Md. Zakaria Hossain and Bujang B.K. Huat －Editors

GEOMATE 2012 Kuala Lumpur
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTiPROCEEDINGS OF SECOND INTERNATIONAL CONFERENCE – GEOMATE 2012
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT KUALA LUMPUR,
MALAYSIA 14-16 NOVEMBER, 2012

Geotechnique, Construction
Materials and Environment
Edited by

Md. Zakaria Hossain
Graduate School of Bioresources
Mie University

Bujang B.K. Huat
School of Graduate Studies
Universiti Putra Malaysia

THE GEOMATE INTERNATIONAL SOCIETYii

Copyright @ 2012 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrity and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: geomate@gi-j.com
www.gi-j.com
ISBN Number: 978-4-9905958-1-4 C3051

iii

Second International Conference on Geotechnique, Construction Materials andEnvironment, Kuala Lumpur, Malaysia, Nov. 14-16, 2012,
ISBN: 978-4-9905958-1-4 C3051

Table of Contents

Prefacexxi

Organization

ID Keynote Papers
1

Use of microbial technology in geotechnical engineering
Jian Chu, Volodymyr Ivanov, Jia He and Viktor Stabnikov

1-4

2

A study on slow moving slopes
L G Tham and Xu Kai

5-11

3

Bearing capacity and settlement behaviour of shallow foundations in unsaturated
sands
Sai K. Vanapalli and Fathi M. O. Mohamed

12-21

4

Development of Piled Geo-wall
Hara, T., Tsuji, S., Yoshida, M. and Sawada, K.

22-29

5

Measurement of strain distribution along precast driven pile during full scale pileload test
Faisal Ali and Lee Sieng Kai

30-39

6

Unique grouting material composed of calcium phosphate compounds
S. Kawasaki and M. Akiyama

40-46

7

Stiffness Anisotropy and Resilient Modulus of an Unsaturated Soil under Static andCyclic loads
Charles W.W. Ng, C. Zhou and J. Xu

47-62

8

Nanoparticles for Geotechnical Engineering
Mohd Raihan Taha

63-68

Technical Papers
Geotechnique24

Damage Assessment and Strengthening of R/C Building Constructed on Expansive
Soils
Osama M. A. Daoud

iv

69

70-75

29

Gravitational Sphere Packing Method To Find The Porosity Of The Soil Particles
With The Identical Size in cylindrical container
Mohammad Mahdi Roozbahani , Bujang B. K. Huat, Afshin Asadi

76-78

34

Parametric Study of Piled Raft Foundations to improve the Design Method
Tohidi A., MH .Sadagiani

79-84

39

Experimental Study of Suction-Monitored CBR test on Sand-Kaolin Clay Mixture
Purwana, Y.M., Nikraz H., and Jitsangiam P.

85-90

41

Foundation Strengthening of Historic Buildings by Micropiles
Azza M. Elleboudy

91-96

44

Group effects of piles due to lateral soil movement
Hongyu Qin1 and Wei Dong Guo

97-10245

Collapse-Settlement Calculations for Embankments Based on a Collapsible Soil
Subgrade
Moshe Livneh

103-108

46

The influence of trench soil properties on the accuracy of pressure cellmeasurements in an earthfill dam
D. Elmi, A. A. Mirghasemi

109-11250

Studying the effects of considering dilation angle over the performance ofgeotechnical elements using numerical methods
Salman Hanifi, Amin Naeimabadi , Mohammad Reza Atrchian113-11750aComparing the excavating effect of single and twin tunnels on the earth surface
settlement by numerical methods
Amin Naeimabadi, Salman Hanifi , Mohammad Reza Atrchian

118-124

107

Excess Pore Pressure Characteristics of Sand Mat using Dredged Soil
M.S LEE, K. ODA

125-130

110

Effect of clays fraction to california bearing ratio laboratory test value with andwithout soaked
Soewignjo Agus NUGROHO, Syawal Satibi and Ferry Fatnanta131-136

111

Compressibility Characteristics of Residual Soil Subjected to Microbial Treatment
Lee Min Lee, Ng Wei Soon, and Yasuo Tanaka

137-140

115

Strength and aging of cement treated low plastic soils
Fabien Szymkiewicz, Antoine Guimond-Barrett, Alain Le Kouby

141-145

119

Performance Of Cantilever Sheet Pile Supporting Excavation Adjacent To Strip
Footing
Yousry M. Mowafe, Ahmed M. Eltohamy and Ahmed A. Mohamed

146-153

119a

Enhancing Stability of Slopes with Cement Kilin Dust Sand Mixture Layer
Al Serif M. Abd Alazez1, Ahmed M. Eltohamy and Ahmed Rusdy Towfek

154-161122

The effect of confining stress on the analysis of excavations with adjacent existingbuildings
Sabzi Zahra, Fakher Ali

162-166

125

Bored Pile Socket in Erratic Phyllite of Tuang Formation
D.E.L. Ong

167-171v

133

Swelling deformation of the bentonite mixed with silica fume or sodium carbonate
in calcium hydroxide solution
Yuki YOKOYAMA and Kenichiro Nakarai

172-175

136

A Combined Probabilistic Method for Engineering Failure analyses andApplications in Typical Geotechnical Structures
Duruo Huang

176-181137

Parametric Study of Crossing Tunnels at Different Level
A. Tohidi, MH.Sadagiani

182-189

143

Use of post-floatation copper tailings in the construction of dump dams
Tschuschke W., Wierzbicki J.

190-195

150

Effects of Slope Inclination on the Rain-induced Instability of Embankment Slopes
Chaminda Gallage, Shiran Jayakody and Taro Uchimura

196-201

225

Extension of an integrated AHP and TOPSIS approach in the earth dam site
selection
Minatour Yasser, Khazaie Jahangir

202-208

227

Using Geogrid Boxes as a New Approach for Reinforcement of Rock Slopes
Ahmad Fahimifar, Arvin Abdolmaleki

209-214228

Volume Change Characteristics of Clay Mixes with Different Bentonite Percentages
J. M. Kate, Sunil Kumar, M. P. Bhorkar

215-218236

FE Analysis for Settlement of Peat Ground under Embankment Loading
Satoshi Nishimoto and Hirochika Hayashi

219-224

237

Seismic behavior of pile group in soil slopes: 1g shake table tests
H. Elahi, M. Moradi,A. Ghalandarzadeh and A. Elahi

225-229

239

Electrophoresis of Bacteria and Electro-Biogrouting
Hamed A. Keykha, Bujang B. K. Huat, Afshin Asadi, Satoru Kawasaki

230-232240

Consideration concerning reasonable modified form of Floating-type Improved
GroundHijiri Hashimoto, Satoshi Nishimoto and Hirochika Hayashi

233-236

241

Numerical Study Effect of Nail Angle Change on Stability and Displacement of
Excavation Wall in Cohesive and Cemented Non-Cohesive Soils
Navid Shahnazi, Jahangir Khazaie, Mohamad Sharifipour and Kiarash Ashtari237-241242

An investigation on behavior of geogrid reinforced soil retaining walls applying
finite difference method (F.D.M)
Jahangir Khazaie and Alireza Sadeghabadi242-252

243

Effects of humic acid and salt additives on the behaviour of lime-stabilised organic
clay.
N.Z. Mohd Yunus, D. Wanatowski, and L.R. Stace

253-256

244

Sodium Tripolyphosphate For Stabilizing Of Contaminated Soil
Amin Falamaki ,Hossein Tavallali and Shahrbanoo Rezanejad Farahmand257-260

246

Evaluation of Soil Water Characteristic Curves for Sand-Expansive Clay Mixture
T. Y. Elkady, M. A. Dafalla, A. M. Al-Mahbashi1 and Mosleh Al Shamrani261-265

vi

247

Coastal soils improvement
Mohammad reza Atrchian, Morteza Ala, Amin Naeimabadi266-269

248

A Hybrid Intelligent System of Fuzzy C-Mean Clustering and Neural Networks to
Predict the Compaction Characteristics of Fine Grained Soils
Iman Ashayeri, Mahnoosh Biglari270-273249

Equivalent linear seismic ground response analysis of unsaturated soil deposits
Mahnoosh Biglari, Iman Ashayeri274-278251

The Correction of Shallow Foundation Settlement Equations by Finite Element
MethodMohammad Arab Ameri, Hakime Rabbanifar and Mehdi Arab Ameri

279-284

255

Numerical study of earthquake hazards of liquefaction occurring on the site andsustainable transport pipes
Hamed Bavanpouri, Jahangir Khazaie, Dr.Hasan Sharafi

285-288

266

Analysis of geogrid reinforced soil retaining walls with limit equilibrium and finite
difference method
Alireza Sadeghabadi ,Jahangir Khazaie and Mostafa Yousefi Rad

289-293

270

Recent Development in Electrokinetic Stabilization of Soft Soil, A review
Hossein Moayedi, Sina Kazemian, and Bujang B. K. Huat

294-298

275

Application of ion-exchange solution soil stabilizer in increasing CBR values ofclayey soils
Reza Ziaie Moayed, Seyed Ali Hashemi and Farzad Allahyari

299-302

276

Assessment of Low Strain Wave Attenuation for Piles Cast in Very Dense Granular
Soil
Ramli Nazir and Osman El Hussien

303-307279

Dynamic Soil-Structure Interaction Considering Pore-Water Pressure Coupling
Qutayba Nazar Al-Saffar, P Akram Y. Al-Sa’aty, Mohamad T. Al-Layla

308-325

281

Analysis and design of tunnel lining in alluvial soil deposits
Nuri al-Mohamadi

326-330285

Effect of silt content on the anisotropic behaviour of silt-sand mixtures
Navid Khayat , Abbas Ghalandarzadeh

331-337

288

Optimizing the twin tunnels position for reduction the surface ground movements
Mehdi Mokhberi, Seyed Ali Farjam

338-344290

Evaluation of pile lateral capacity in clay applying evolutionary approach
Alkroosh I., Nikraz H

345-348

293

Effect of Reservoir Drawdown Rate on Phreatic Line Recession in Homogeneous
Earth Dams
R. Ziaie Moayed, V. Rashidian349-352294

The effect of lime-microsilica stabilization on California Bearing Ratio of silty soils
R. Ziaie Moayed, Y. Daghigh2, B. Pourhadi and N. Sahebzamani

353-356295

Effect of elastic modulus varieties in depth on modulus of subgrade reaction on
granular soils
Reza Ziaie Moayed, Mahdi Ali Bolandi357-360vii

296

Flexural Buckling Behavior of Steel Pile with Vertical Load in Liquefied Soil
Yoshihiro Kimura and Koichi Onohara

361-365

305

Two-dimensional finite difference pseudo-static deep-seated slope stability analysis
of embankments over stone column-improved soft clay
Farshad Fayyaz Jahani, Mohammad Sharifipour

366-371

308

Parametric study of Geo-grid soil reinforced retaining walls under dynamic loading
by ABAQUS software
Atrchian mohammad reza, Daghigh younes, Ghanbari namin hamed372-3802010

Sensitivity analysis in soil nail wall for Tehran-Tabriz railway by considering thecoefficient of variation for effective parameters in stability analysis
Mohammad Moghadaripour , Ardalan Akbari Hamed, Ali Ghozat

381-3842012

Evaluation of Strength Parameters of Improved Collapsible Soils with Current
Additives (Case study: Atrak zone, Iran)
Mohammad Arab Ameri, Samaneh Kargar and Mehdi Arab Ameri

385-390

2100

Influences of soil material parameters in compaction simulation with soil/water/air
coupled F.E. code
S. Sakamoto, K. Kawai, V. Phommachanh and A. Iizuka

391-3962105

Lower bound limit analysis of strip footings resting on cohesive soils
S.M. Binesh, A. Gholampour

397-4022106

Upper bound limit analysis of strip footings resting on cohesive soils
S.M. Binesh, S. Raei

403-4082108

Evaluating and optimizing different methods of treating and cleaning contaminatedsoil
Mohammad Reza Atrechian,Seyed Salman Musavi Khosro Ebrahimkhani

409-412

2112

Experimental study concerning impact characteristics by collision of weight onsand cushion over steel beam
Tam Sy Ho, Hiroshi Masuya and Naoto Takashita

413-4192114

Landslides in Tea Plantation Fields in Shizuoka, JapanJun Sugawara

420-425

2115

Natural base isolation system for earthquake protection
Srijit Bandyopadhyay, Aniruddha Sengupta and G.R. Reddy

426-4312116

Comparison of Effect of Percent of Compaction, Sand and Cement on The Amountand Time of Swelling of Bentonite
Mohammad S. Pakbaz and B. Rafiee

432-435

2118

Prediction of the Axial Bearing Capacity of Piles by SPT-based and Numerical
Design Methods
Issa Shooshpasha, Ali Hasanzadeh and Abbasali Taghavi436-4392120

Application of recently developed mechanistic-empirical methods for the design ofheavily loaded pavement structures
John N. Mukabi

440-447

2121

Design of support system for excavation in black cotton soils in Guntur, India
D. Neelima Satyam, Akhila Manne

448-451

2127

Numerical modeling of pullout capacity of a pile in sand under oblique load
Md. Iftekharuzzaman and Dr. Bipul Hawlader452-457

viii

2130

Slope Failure in Residual Soils of Peru
A.Carrillo-Gil

458-461

2131

Back-analysis of Frictional Jacking Forces Based on Shear Box Testing of
Excavated Spoils
Choo Chung Siung and Dominic Ong E. L.

462-467

2132

Lime Stabilization of Tropical Soft Soils
Dygku Salma Awg Ismail, Tan Jui Siang and Siti Noor Linda Taib468-471

2138

Increase Of Piles Bundle Capacity By Making A Base Joining Them
Jan Jaremski

472-478

2139

The Effects of soil-Pile Interaction on Seismic Parameters of Superstructure
Mahdy Khari1, Khairul Anuar Bin Kassim, Azlan Bin Adnan

479-4842140

Investigation on Finite Element Modeling of Group of Stone Columns
M. Mirshekari , C. Behnia and S. Fakhretaha

485-4882147

Suction Controlled Triaxial Apparatus for Saturated-Unsaturated Soil Test
Luky Handoko, Noriyuki Yasufuku, Kiyoshi Oomine, Hemanta Hazarika

489-4932154

Monitoring Program and Dilapidation Survey for Grouting of Karst Cavities
Hasan A. Kamal and Hassan J. Karam

494-4972156

Numerical Analysis of Soil Nail Walls under Seismic Condition in 3D Form
Excavations
Siavash Zamiran, Hamidreza Saba

498-501

2159

Shear strength improvement of fibrous peat due to consolidation pressure
Vivi Anggraini, Nurly Gofar and Bujang bin Kim Huat

502-506

2164

Factors affecting the stability analysis of earth dam slopes subjected to reservoir
drawdown
T. Souliyavong, C. Gallage, P. Egodawatta and B. Maher

507-5122165

Lateral stress induced due root-water-uptake in unsaturated soils
Mu’azu Mohammed Abdullahi and Nazri Bin Ali

513-5172166

The effect of pre-tensioned rock bolts on seismic behavior of urban tunnels
Farshad Fayyaz Jahani, Mohammad Sharifipour

518-5232170

Effect of silt content on the anisotropic behaviour of silt-sand mixtures
Navid Khayat , Abbas Ghalandarzadeh

524-530

Construction Materials
24b

Production and Properties of High Strength Concrete for Concrete Dam
Heightening Project in Sudan
Osama M. A. Daoud and, H. S. Sagady

531-537

27

Effect of openings locations of quadratic folded plate roofing systems on their staticand dynamic behavior
Ahmed Hassan

538-544ix

48

Improving concrete compression strength using natural additive
Abdoullah Namdar, Ideris Bin Zakaria and Nurmunira Binti Muhammad Atan

545-548

53

DEM simulations and laboratory experiments on physical and mechanicalproperties of sand-gravel mixtures
Janaka J. Kumara, Kimitoshi Hayano, Yuuki Shigekuni and Kota Sasaki

549-554

120

Rock slope stability problems in Gold Coast area in Australia
Shokouhi Ali, Gratchev Ivan, Charrismanagara Arry

555-558

121

Techno-commercial development of pre-fabricated structures using geopolymer
concrete
J. Santhosh, R.Sanjeev, P.Sethu Raja, A. Rajendra Prasad, A. Abdul Rahman

559-563

124

Mechanical Properties of Recycled Steel Fibre Reinforced Concrete
A.S.M. Abdul Awal, Lim Lion Yee, M. Dianah1 and M. Zakaria Hossain

564-569

134

Effects of Gradation of Various Sand Deposits of Pakistan on Strength of Hardened
Concrete
Gauhar Sabih , S. Muhammad Jamil and Kamran Akhtar

570-574

203

Assessment of recycled concrete aggregates for road base and sub-base
S. Jayakody, C. Gallage, A.Kumar

575-579232

Estimation of consolidation properties of Holocene clays with artificial neuralnetwork
Kazuhiro ODA , Minsun Lee and Shotaro Kitamura

580-585250

Measurement of The Crack Displacement Using Digital Photogrammetry for
Evaluation of The Soundness of Tunnels
A. Kanazawa, S.Nishiyama, T.Yano, T.Kikuchi

586-592

259

Numerical modelling of a small scale shallow foundation reinforced by
Soil-Mixing
Anna GRZYB, Mahmoud Dhaybi and Frederic Pellet

593-598

262

Physical modelling of a small scale shallow foundation reinforced by Soil-Mixing
Mahmoud Dhaybi and Frederic Pellet

599-604282

Prediction compressive strength of concretes containing silica fume andstyrene-butadiene rubber (SBR) with a mathematical model
A.Haghollahi, M.Shafieyzadeh

605-610283

287

Sliding stability of dry masonry block retaining structure with a resistance plate
Akihiro Hashimoto, Noriyuki Yasufuku, Yoshio Suematsu, Kazuo Fujita1 andToshimitsu Komatsu
Evaluation and Prediction Method on Neutralization of Supplied Long-term
Hydraulic Concrete Structure
Man-Kwon Choi , Yuki Hasegawa, Shinsuke Matsumoto, Shushi Sato and Tsuguhiro
Nonaka611-616617-621297

Measurement of tensile properties of geogrids
Raid R. Al-Omariand Mohammed K. Fekheraldin

622-631306

Modeling the Swelling strain and pressure of Weak Rock using Adaptive
Network-Based Fuzzy Inference System
Ramin Doostmohammadi

632-636309

Centrifuge Modeling Of Reinforced Embankments On Soft Foundation
Ali Sobhanmanesh, Nurly Gofar

637-642x

2102

Centrifuge model tests and finite element analyses on the seismic behavior of thequay walls backfilled with cement treated granulate soils
Hayano, K., Morikawa, Y., Fukawa, H., Takehana, K. and Tanaka, S.

643-6502109

Response of Piered Retaining Walls to Lateral Soil Movement Based on Numerical
Modeling using Matlab
Siamak Pahlevanzadeh

651-6542122

Evaluation of Thermal Response Tests in Energy Piles
Seung-Rae Lee, Seok Yoon, Hyun-Ku Park , Gi-Dae Oh

655-659

2134

Experimental Study on Mechanical Characteristics of Soil-Geosynthetic Interface
Md. Bellal Hossain, Md. Zakaria Hossain and Toshinori Sakai

660-665

2136

An Interpretation of Mechanical Properties of Bentonite as a Non-linear Elastic
Material
Y. Takayama, S. Tsurumi, A. Iizuka, K. Kawai

666-6712149

Effect of carbonation on strength of cement treated sandKenichiro Nakarai and Tomomi Yoshida672-675

2167

Application of PML to Analysis of Nonlinear Soil-Structure-Fluid Problem using
mixed element
Pahaiti Reheman, Hiroo Shiojiri

676-681

Environment

64

28

Recycled Bassanite as a Stabilizing Agent for Cohesion-less Soil
Aly Ahmed and Keizo Ugai

682-687

31

The effects of environmental on today construction industry
Mahdy Khari1, Khairul Anuar Bin Kassim, Azlan Bin Adnan

688-692

33

A Case History of Quay Wall Failure and Remedial Design
C. S. Chen

693-696

37

Environmental economical efficiency in treatment/reutilization of construction
sludges considering delay in process flow
Shinya Inazumi, Hiroyasu Ohtsu and Takayuki Isoda

697-70242

Evaluation and validation of elastic-fragile damage modeling for unsaturated
porous media in θ-stock
B. Gatmiri , M. Fathalikhani

703-708

43

47

Estimation of the Landslide Dam Sustainability in the Gigantic Seimareh Landslide
Using the Rate of Sedimentation
Zieaoddin Shoaei
Influence of soil and rock mineralogy and geomorphology on Landslides
Occurrence and soil erosion in the Merek Catchment, IranMosayeb Heshmati , Nik Muhamad Majid, J. Shamshuddin, Arifin Abdu andMuhamad Ghaituri709-7012

713-719112

The effect of historical land use on landslide initiation
Gholamreza Shoaei

720-723

114

Integrated Environmental Management for Sustained Development
J.Rajaraman and K.Thiruvenkatasamy

724-729xi

114a

Lowland Environmental Geotechnology of Seismosediments of Kandla Port In
India
J.Rajaraman and K.Thiruvenkatasamy

730-735

123

Single step extraction to find out soluble lead in soil
Masahiko Katoh, Satoshi Masaki, Takeshi Sato736-741

127

Biogas Production from Tannery Sludge
A.Rajendra Prasad, A.Dhanalakshmi, S.Usha and A.Abdul Rahman

742-746

131

Water Harvesting and Salinization Prevention by Capillary Barrier of Soil
Toshihiro Morii, Mitsuhiro Inoue, Kaoru Kobayashi, Tetsuya Suzuki and Takayuki
Kawai

747-752

204

Study of the Performance of anaerobic digestion of greasy skim as a renewableenergy source
Salam J. Bash Al-Maliky

753-756

209

Seismic evaluation of NARGES soil nailed wall under Cyclic loading and Pseudo
Static forces
Ardalan Akbari Hamed , Mohammad Moghadaripour , Ali Ghozat

757-761229

Purification System of Ocean Sludge by and Using Coagulants and then Activating
Microorganisms
Kyoichi Okamoto and Kenji Hotta

762-767

233

Mechanical properties of volcanic products mixed with industrial wastes
K. Yamamoto, T. Negami2, M. Hira, N. Aramaki and Y. Hayashi

768-771

235

Measurement of Hydraulic Conductivity for Peat Ground Using CPTU
Hirochika Hayashi and Satoshi Nishimoto

772-775

252

Extraction of temporal and spatial properties on habitat of snail by means ofstatistical approach
Masaaki KONDO and Takamitsu KAJISA

776-781

265

267

268

271

274

280

286

The use of Nano Zero Valent Iron to remediation of contaminated soil andgroundwater
Taghizadeh Maryam , Yousefi Kebria Daryoush and Gholamreza Darvishi
A Multi-parameter Water Quality Analysis by onsite filtered 2mL Sample to
Monitor Urban River Eutrophication
Akira Kikuchi, Nor Eman Ismail, Narges Janalizadeh, Musa Mutah and Muhamad
Faiz
The intelligent performance installations in energy efficiency in green buildings
Hooman Abadi, Rouzbeh Abadi, Poona Abadi
Study the Effect of Polyvinyl Chloride (PVC) Addition on the Permeability ofAnbar Soils
Ahmed H. Abdul Kareem & Omar Mustafa
Monitoring deforestation and rangeland destruction through land use alteration
during 1955- 2002, using GIS and RS in Merek sub-basin, west Iran
Mohammad Gheitury, Mosayeb Heshmati, Mohammad Ahmadi and Nik Muhamad
Majid
The comparison of phytoremediation and electrokinetic methods in remediation of
petroleum hydrocarbons contaminated soil
Gholami Meade, Yousefi Kebria Daryoush
Using Metacognitive Strategies On Listening Comprehension
Mahshid Mirzaaghaeexii

782-785786-790791-796797-802803-807808-811812-822

298

Evaluation of Run-Off Supply Projects in Hamedan Province(Iran)N. Rostam Afshar, M. Abdoli

823-827

299

Protective measures of the monumental pine in Rikuzentakada withsoil-water-air-solved material coupled model
Y. Sugiyama, S. Nomura, K. Kawai and A. Iizuka

828-833

304

Effects of Biosurfactants in Electrokinetic Remediation of contaminated soils
Bour Moslem and Yousefi kebria Daryoush834-838

2117

Change in structure of ground-beetle assemblage on river areas after construction of
biotopes using the technique, “Nature oriented river works”
Michko Masuda, Takaaki Tsukada and Fumitake Nishimura

839-8442123

Dense granular-fluid mixture: the effects of the natural grains characteristics andthe interstitial fluid viscosity
Anna Maria Pellegrino and Leonardo Schippa

845-850

2131

Back-analysis of Frictional Jacking Forces Based on Shear Box Testing of
Excavated Spoils
Choo Chung Siung and Dominic Ong E. L.

851-8562137

An idea of multi-functional reservoirs in mountain regions
Jan Jaremski

857-8622141

Estimation Method of Amount of Tsunami Disaster Wastes during the 2011 off the
Pacific Coast of Tohoku Earthquake
Minoru Yamanaka, Naoya Toyota, and Shuichi Hasegawa, Atsuko Nonomura

863-866

2143

Assessment of Runoff in the High Humid Foot-hill Areas of Arunachal Himalayas
Using Thornthwaite Equation
Md. Eahya Al Huda and Surendra Singh

867-871

2161

2163

2171

Natural Radiation level in Groundwater from Katsina, a Semi Arid Region of
Northern Nigeria
B.G. Muhammad and M.S. Jaafar
Cost-Performance Studies of Abandoned Shell Husks (ASH) for Soil
Reinforcement Applications
Md. Toriqul Islam, Md. Zakaria Hossain, Masaaki Ishida, Md. Bellal Hossain
and Md, Jamal Uddin
Cyclones Path and Severities: Effect on Flood Height and Damages of Land and
Embankments in BangladeshMusahaq Ali, Zakaria Hossain and Shigeko Haruyama
Author index

872-876877-880881-884885

xiii

Preface

Second International Conference on Geotechnique, Construction Materials
and Environment, Kuala Lumpur, Malaysia, Nov. 14-16, 2012,
ISBN: 978-4-9905958-1-4 C3051

The Second International Conference on Geotechnique, Construction Materials and Environment
GEOMATE2012, held at the Istana Hotel, Kuala Lumpur, Malaysia, was organized by the School ofGraduate Studies, Universiti Putra Malaysia, in collaboration with the GEOMATE International Society,
AOI-Engineering, the Useful Plant Spread Society, HOJUN and Glorious International. Its aim was toprovide scholars the space and opportunity to share common interests on geo-engineering, construction
materials, environmental issues, water resources, and earthquake and tsunami disasters. The key objective of
this conference was to promote interdisciplinary research from different regions across the globe.
On Friday 11 March 2011, at 14:46 Japan Standard Time, the north east of Japan was struck and severely
damaged by a series of powerful earthquakes which also caused a major tsunami. This conference wasdedicated to the tragic victims of the Tohoku-Kanto earthquake and tsunami disasters. This was the second
event after the first international conference that was successfully held in Tsu city, Mie, Japan, with morethan 150 participants.
The conference had three major themes with 17 specific themes including:
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Advances in Composite Materials
Computational Mechanics
Foundation and Retaining Walls
Slope Stability
Soil Dynamics
Soil-Structure Interaction
Pavement Technology
Tunnels and Anchors
Site Investigation and Rehabilitation
Ecology and Land Development
Water Resources Planning
Environmental Management
Public Health and Rehabilitation
Earthquake and Tsunami Issues
Safety and Reliability
Geo-Hazard Mitigation
Case History and Practical Experience

The proceedings from this conference contain eight keynote papers along with 145 technical papers from 25
countries. The technical papers were selected from the vast number of contributions submitted, after a reviewof the abstracts. The final papers selected to be included in the proceedings have been peer reviewed
rigorously and revised as necessary by the authors. We are especially grateful to the authors for theirresourcefulness and their help in maintaining the high standard of the collection, as well as their co-operationin complying with the requirements of the editor and the reviewers. We wish to express our sincere thanks to
the Organizing Committee Members, the National Advisory Committee Members and the International
Advisory Committee Members for their invaluable support. We would also like to acknowledge the supportof the School of Graduate Studies, Universiti Putra Malaysia, the GEOMATE International Society,
AOI-Engineering, the Useful Plant Spread Society, HOJUN and Glorious International.

Prof. Dr. Bujang B.K. Huat
Conference Chairman

Dr. Md. Zakaria Hossain
Conference SecretaryxivSecond International Conference on Geotechnique, Construction Materials
and Environment, Kuala Lumpur, Malaysia, Nov. 14-16, 2012,
ISBN: 978-4-9905958-1-4 C3051

Organization
Scientific Committees:
Conference Chairman: Prof. Dr. Bujang B.K. Huat, Universiti Putra Malaysia, Malaysia
Conference Honorary Chairman: Emeritus Prof. Dr. Sohji Inoue, Mie University, Japan
Conference Organizing Committee:
Prof. Faisal Ali, Universiti Pertahanan Nasional Malaysia
Prof. Mohd Raihan, Taha Universiti Kebangsaan Malaysia
Prof. Roslan Hashim, Universiti Malaya
Prof. Thamer A. Mohamed, Universiti Putra Malaysia
Prof. Dato' Ismail Bakar, Universiti Tun Hussein Onn Malaysia
National Advisory Committee:
Prof. Dr. A.S.M. Abdul Awal, Universiti Technologi Malaysia, Malaysia
Prof. Dr. Ideris Bin Zakaria, Univ. Malaysia Pahang, Malaysia
Prof. Dr. Muhammad Ibn Ibrahimy, International Islamic University, Malaysia
A/Prof. Dr. Zunaidi Bin Ibrahim, Deputy Reactor, Kolej Universiti TATI, Malaysia
A/Prof. Dr. Abdoullah Namdar, Universiti Malaysia Pahang, Malaysia
Prof. Dr. Suleyman Aremu Muyibi, International Islamic University, Malaysia
Prof. Dr. Shahjahan Mridha, International Islamic Univ., Malaysia
Prof. Dr. Faisal Hj Ali, National Defense University of Malaysia
International Advisory Committee:
Prof. Dr. Fumio Tatsuoka, Tokyo University of Science, Japan
Prof. Sai Vanapalli, University of Ottawa, Canada
Prof. Dr. Nemy Banthia, University of British Columbia, Canada
Prof. Dr. Ian Jefferson, University of Birmingham, UK
Prof. Dr. John Bolander, Univesity of California, USA
Prof. Dr. Shamshul Chowdhury, Roosevelt University, USA
Prof. Dr. Isabel Pinto, University of Coimbra, Portugal
A/Prof. Mark Jaksa, University of Adelaide, Australia
A/Prof. Dr. William Sebudde Kaggwa, University of Adelaide, Australia
Dr. John N. Mukabi, President of Kenya Geotechncal Society, Kenya
Prof. Dr. Jing-Cai Jiang, University of Tokushima, Japan
A/Prof. Dr. Kingshuk Roy, Nihon University, Japan
A/Prof. Dr. Md. Shahin Hossain, Nagoya Institute of Technology, Japan
Prof. Dr. Mounir Bouassida, National School of Engineering of Tunis
Prof. Dr. Lilia Robles Austriaco, Angles University Foundation, Philippines
Prof. Dr. Zahurul Islam, Bangladesh University of Engineering and Technology, Bangladesh

xvConference Secretariat:
Dr. Zakaria Hossain, Associate Professor
Division of Environmental Science and Technology
Graduate School of Bioresources
Mie University, 1577 Kurima Machiya-cho
Tsu-city, Mie 514-8507, Japan
E-mail: zakaria@bio.mie-u.ac.jp
Tel: +81-59-231-9578、Fax: +81-59-231-9591
Editorial Committee:
Md. Zakaria Hossain
Prof. Dr. Bujang B.K. Huat
Executive Committee:
Bujang B.K. Huat
Md. Zakaria Hossain
Muhammad Ibn Ibrahimy
Md. Bellal Hossain
Afshin Asadi

xvi

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Use of Microbial Technology in Geotechnical Engineering
Jian Chu1,2, Volodymyr Ivanov2, Jia He2 and Viktor Stabnikov2
1 Department of Civil, Construction & Environmental Engineering, Iowa State University, USA
2 School of Civil and Environmental Engineering, Nanyang Technological University, Singapore
ABSTRACTThe recent development and principles of the application of microbiology in geotechnical engineering are briefly
introduced. The microbial approach has several advantages over the conventional soil improvement methods. It is morecost effective, energy saving, environmentally friendly, and simpler to implement. Three types of applications that havebeen studied recently are introduced. The first is improvement of shear strength of soils by introducing crystallizations
or slimes using the microbial process. The second is mitigation of liquefaction potential of soils by forming uniformly
distributed and stable gas bubbles in liquefiable soil layers using bacteria. The third is seepage and erosion control byforming a hard layer of calcite crust on the soil surface using the microbial urea hydrolyzing process.
Keywords: geotechnical engineering, grouting, microbiology, soil improvement

penetrate and reproduce themselves in soil. Harnessing thisnatural, unexhausted resource may result in an entirely new
approach to geotechnical or environmental engineering
problems and bring in enormous economic benefit toconstruction industries. The application of microbial
biotechnology to construction will also simplify some of theexisting construction processes. For example, the biocement
can be in either solid or liquid form. In liquid form, thebiogrout has much lower viscosity and can flow like water.
Thus, the delivery of biocement into soil is much easiercompared with that of cement or chemicals. Furthermore,when cement is used, one has to wait for 28 days for the fullstrength to be developed, whereas when biocement is used,
the reaction time can be much reduced if required.

1. INTRODUCTION
When there are constructions over soft ground, the soft soil
needs to be treated. There are a number of ways to strengthen
soft or weak soil. One of the common approaches is to use
cement or chemicals to increase the load bearing capacity or
the so-called shear strength of soil. The same process can beused to reduce the water conductivity of soil or the rate ofwater flow in soil. This is necessary when there is a need toprevent water from flowing in the ground, for example, for
cutting off contaminated groundwater flow. In the above
cases, cement or chemicals are used as a binder to mix withsoil to either increase the shear strength or reduce the water
conductivity of soil. However, the use of cement or chemicals
for construction or soil improvement is not sustainable in thelong run as cement or chemicals require a considerable
amount of natural resource (for example limestone) andenergy to produce. The production process also generates
carbon dioxide, dust and possibly other toxic substances andthus is not environmentally friendly. The use of cement or
chemicals for soil improvement is also expensive and time
consuming. There is an urgent need to develop new andsustainable construction materials that can reduce the use of
cement or chemicals for geotechnical applications.
Using the latest microbial biotechnology, a new type of
construction material, biocement, has been developed as analternative to cement or chemicals [1], [2]. Biocement ismade of naturally occurring microorganisms at ambient
temperature and thus requires much less energy to produce. It
is sustainable as microorganisms are abundant in nature andcan be reproduced easily at low cost. The microorganismsthat are suitable for making biocement are non-pathogenic
and environmentally friendly. Furthermore, unlike the use of
cement, soils can even be treated or improved without
disturbing the ground or environment as microorganisms can

2. PRINCIPLES
The principles of microbial treatment are to use themicrobially-induced calcium carbonate precipitation or otherapproaches to produce bonding and cementation in soil so as
to increase the strength and reduce the water conductivity ofsoil. A number of studies have been carried out in recent years[2]-[5]. Much of the work still stays at the experimental stage.
However, the scale of treatment has increased rapidly withtime and has reached 100 m3 in the recent years [5].
The microbiological processes induce calcium carbonate
crystals, other minerals or slimes as illustrated by examples
shown by Van der Ruyt and van der Zon (2009) [4], Van
Paassen et al. (2010) [5] and Chu et al. (2012) [6]. Thoseminerals or slimes act as cementing agencies between sandgrains to increase the shear strength of soil and/or to fill in the
pores in soil to reduce the water conductivity as illustrated
schematically in Fig. 1. The two processes to increase
strength and reduce conductivity have been calledbiocementation and bioclogging respectively [2]. The process1to deliver the biocement in-situ to achieve biocementation or
bioclogging is called biogrouting. As the viscosity of
biogrout is low, it is possible to pump in the biogrout into theground without mixing for sandy soil. This will enable theconstruction process to be simplified. The existing study so
far shows that the biocement method is effective in bothincreasing the shear strength and reducing the water
conductivity of soil.

Fig. 2 as an example. When cement or chemicals are used totreat soil, the amount of improvement in the shear strength ofsoil is dependent on the amount of cement or chemical used.
Similarly, when biocement is used, the shear strength of soil
is affected by the amount of metal precipitation. One way tomeasure the shear strength of soil is by simply compressing asoil column in between of two rigid plates, the so-called
uniaxial compression test. The shear strength measured bythis method is called the uniaxial compressive strength(UCS). In one study by Van der Ruyt and van der Zon (2009)[4], the UCS of biocement treated sand was measured forspecimens having different calcium carbonate contents. The
results are shown in Fig. 3. It can be seen that the UCS
strength increases with increasing calcium carbonate content.
The highest UCS obtained is 27 MPa. For normal
applications, the UCS strength required is less than 3 MPa.
This will only require a calcium content of 100 to 200 kg/m3.
To achieve the same UCS strength for sand using cement
grounting, the amount of cement used would be between 250
to 300 kg/m3. As the production of biocement can be cheaper
as discussed by Ivanov and Chu (2008) [2], the overall costfor biogrouting can be potentially lower. Another advantage
is that when cement grouting, it takes 28 days for the UCS toreach the targeted value. However, when biocement is used,full strength can be gained in a much shorter time if required.

Fig. 1. Schematic Illustration of biocementation and bioclooging
processFig. 3. The unconfined compression strength (UCS) versus
calcium carbonate content relationship for biocement treatedsand (after Van der Ruyt and van der Zon, 2009)

3.2 Mitigation of Soil Liquefaction by Biogas
Sol liquefaction refers to a phenomenon in which a soil istransformed into a substance that acts like a liquid in responseto an external action such as earthquake. When liquefaction
occurs, the ground losses completely it’s bearing capacity andundergoes large deformation. Soil liquefaction normallyoccurs in saturated sand deposits during earthquake. The
ground shaking will cause the water pressure in soil or theso-called pore water pressure to build up. When the porewater pressure has increased to a certain point, soilliquefaction will occur. Soil liquefaction has been one of the
major causes for serious damages and earthquake related
disasters. More recently liquefaction was largely responsible
for extensive damage to residential properties in the eastern
suburbs and satellite townships of Christchurch, New
Zealand during the 2010 Canterbury earthquake and more
extensively again following the Christchurch earthquakes that
followed in early and middle 2011.

Fig. 2. Sand column treated by biocement

3. POTENTIAL APPLICATIONS OF BIOCEMNT
3.1 Enhancing Shear Strength of Sand
By using the microbially-induced calcium carbonate
precipitation method, the shear strength of soil can beincreased. We know from our childhood experience in castle
building on the beach that dry sand will not stand. However,when dry sand is treated by biocement, a sand column not
only can stand, but also can sustain a lot of weight as shown in
2

The common methods that can be adopted for mitigation ofsoil liquefaction include densification and groundmodification using cement or chemicals. A new approachthat is being developed is the so-called biogas method. In thismethod, tinny gas bubbles are generated in-situ in saturated
sand at where liquefaction may occur. When saturated sand ismade slightly unsaturated by the inclusion of gas bubbles, theamount of reduction in the pore water pressure generated insand under a dynamic load will be greatly reduced. Based on
our research, if we replace only about 5% of water by gas byvolume, we will be able to increase the liquefaction resistanceof loose sand by more than 2 times.
It is not easy to introduce gas into ground. Pumping can beused. Fig. 4 shows a schematic illustration of how to carry out
air pumping for liquefaction mitigation. However, thedistribution of gas bubbles introduced by pumping will not beeven. Furthermore, the gas pumped into ground tends topresent in the form aggregated gas pockets rather than
individual bubbles. As a result, the gas tends to escape fromthe ground. One of the most effective ways to introduce tinnygas bubbles in-situ is to use microorganisms. This method has
the following three advantages over the existing methods: (1)Biocement is like water in the liquid form and flow easily insand. Gases can be generated easily by bacteria anywhere
underground by consuming only a small amount of energy.
Thus the biogas method will be much more cost-effective
than any other methods. As the scale of treatment forliquefaction is normally very large, the potential economic
benefit is significant; (2) The gas bubbles generated bybacteria can be distributed more evenly than other means.
This is because biocement can be delivered by water flow insand and the gas bubbles are generated in-situ rather thanpumped; (3) The gas bubbles generated by bacteria can bemuch smaller. It is necessary for the gas bubbles to be tinnier
so the gas bubbles are less prone to escaping from the ground.

Compared with conventional soil liquefaction mitigation
methods, the biogas method has some significant advantages.
Conventional methods, such as compaction, cementation,
lowering ground water table, and pore pressure relief using
vertical drains, etc., are usually expensive when applied tolarge areas, or need high maintenance. However, the biogas
method is rather cost effective. One reason is that density ofgas is very small, about three orders of magnitude smaller
than solid and liquid, so relatively small amount of solid or
liquid substrates can produce relatively large volume of gas.
The other reason is that the operation is easy, because theliquid phase substrates can easily spread throughout the sandlayer, thus the construction fee can be largely reduced. Other
advantages of the biogas method include environmentallyfriendly, energy saving, and time saving. These advantages
make the biogas method a very promising way to solveliquefaction-related problems in large areas, such as lateral
spreading.

Fig. 5. Comparison of ground settlement induced by groundshaking under an acceleration of 1.5 m/s2 for a saturated sandlayer and a sand layer with 5% gas replacement.

3.3 Seepage and Erosion Control
Biocement can also be used to reduce the water conductivityof sand through the so-called bioclogging mechanism asshown in Fig. 1. One of the methods that has been developedby our research group is to use urea reducing bacteria toprecipitate a layer of calcium carbonate on top of sand as
shown in Fig. 6. This hard layer of crust has a water
conductivity of less than 10-7 m/s and thus can be used as animpervious layer for water storage or for erosion control ofbeach or riverbank. As the layer of treatment is rather thin, theamout of biogrout used is small. Thus the method can be more
economical than convertional methods. The detail of thismethod is described in Chu et al. (2012) [6].

Fig. 4. Mitigation of soil liquefaction by air injection

Some model tests using a laminate box and a shake table togenerate ground motion were carried out. A comparison ofground settlement for a fully saturated sand layer and a sandlayer treated with biogas is made in Fig. 5. The settlement isexpressed as a settlement ratio with the settlement for fully
saturated sand as 100%. It can be seen from Fig. 5 that with
only 5% of gas replacement, the ground settlement generated
under ground shaking with an acceleration of 1.5 m/s2 can be
reduced by more than 90%. Thus, the biogas method iseffective in preventing the occurrence of soil liquefaction or
reducing the damage caused by liquefaction.

4. CONCLUSIONS
The recent development in the application of microbiology in
geotechnical engineering was briefly reviewed in this paper.
The biogrouting method has a number of advantages over theconventional grouting methods. Several potential3

applications were discussed. These include enhancing theshear strength of sand, mitigation of liquefaction of sandusing biogas and control of seepage and erosion of sand by
bioclogging. The biogrouting method paves a morecost-effective, sustainable and environmentally friendly
solution to ground improvement. However, a lot of researchstudies are still required to be carried out before this new
approach can be developed into a common practice.

biomediated ground improvement by ureolysis:
Large-scale biogrout experiment" ASCE Journal of
Geotechnical & Geoenvironmental Engineering, vol.
136(12), 2010, 1721-1728.
[6] Chu J, Stabnikov V and Ivanov V, “Microbially induced
calcium carbonate precipitation on surface or in the bulkof soil”, Geomicrobiology Journal, vol. 29, 2012,
544–549.

(a)(b)
Fig. 6. Formation of (a) a thin impervious layer on top of sandusing a biogrouting method and (b) a water pond model in sandusing this method

5. REFERENCES[1] Whiffin VS, “Microbial CaCO 3 precipitation for the
production of biocement”, Ph.D. Thesis, Murdoch
University, 2004.
[2] Ivanov V and Chu J, “Applications of microorganisms togeotechnical engineering for bioclogging andbiocementation of soil in situ.” Reviews inEnvironmental Science and Biotechnology vol. 7(2),2008, 139-153.
[3] Mitchell JK and Santamarina JC, “Biologicalconsiderations in geotechnical engineering.” ASCE
Journal of Geotechnical and Geoenvironmental
Engineering, vol. 131(19), 2005, 1222-1233.
[4] van der Ruyt M, & Van der Zon W, “Biological in situ
reinforcement of sand in near-shore areas.” GeotechnicalEngineering, vol. 162(1), 2009, 81-83.
[5] van Paassen LA, Ghose R, van der Linden TJM, van derStar WRL and van Loosdrecht MCM, "Quantifying
4Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

A study on slow moving slopes
L G Tham and Xu Kai
Department of Civil Engineering, The University of Hong Kong
ABSTRACTThis paper investigates the mechanism of slow moving slopes by carrying out full scaled field monitoring. Two typicalslopes, representing engineered as well as natural slopes, were studied. In the field monitoring, the variation of porewater pressure, soil moisture content, deformations were measured for more than a year so that one can establish the
behaviour of the slopes for both the wet and dry seasons. Based on the monitored results of the hyrdogeologicalresponses, hydrogeological models were proposed. The subsurface deformations recorded were analyzed and thedeformation mode was also discussed. Furthermore, laboratory tests were carried out to determine the soil properties(strength, stiffness, permeability and soil water characteristics curve) for the subsequent numerical modelling, which
was based on a coupled seepage and deformation finite element model. The numerical results were compared with thefield monitored data. The study finds that the deformation of the slopes is closely associated with the change in the
groundwater table : the slopes will move downhill during rain but they may rebound during the dry period after rain.
The re-bounce could be considerable during the dry season. Nevertheless, the net movement is still downhill annually.
Keywords: Slope, movement, numerical simulation, rainfall• The monitored results indicate that the slope apparently
exhibits continuous cycles of uphill and downhill surface
movements. For this site, the rate of net surface downhill
movement is approximately 1.5 mm/year. The movements
are more pronounced within the colluvium layer andcompletely decomposed tuff (CDT) layers (i.e. the top 2 m
to 4 m), and are above a set of pre-existing slip surfacesthat were identified by ground investigations in 2001.
• The piezometer records reveal that the main groundwatertable is located at about 10 m to 12 m below groundsurface. Seasonal fluctuations caused the water level tovary within ± 5 m. The Halcrow bucket measurements
indicate that the perched groundwater level may be attimes as high as 0.5 m below ground surface.
In general, the slope body will move downslope. However, it
is inconclusive whether the slope will rebound during dry
season or not. In order to have a better understanding on theslow moving slopes under the influence of seasonal
wetting-drying cycles, two slopes (Kwong and Leung[21,22] and Xu [23]) were investigated by carrying out field
monitoring and numerical simulation. This paper will
highlight the findings of the study.

1. INTRODUCTION
Rainfall is reckoned as one of the main triggering factors oflandslides. Rainfall infiltration into unsaturated soil slopes(e.g., Li et al [1]; Zhan et al [2]; Olivares and Damiano [3]),and the influence of infiltrated water on soil suction and shear
strength (e.g., Melinda et al [4]; Farooq et al [5]; Collins andZnidarcic [6]; Zhang et al [7]) have been investigated byexperimental or numerical approaches. Take and Bolton [8]conducted centrifuge tests on an over-consolidated clay
embankment subjected to seasonal weather cycles. They
reported seasonal fluctuation in pore water pressure initiatedprogressive failure. During the wet seasons, the reduction insoil suction induced down-slope movements but the rate wasgreatly reduced during the subsequent dry seasons. Some ofthe downhill deformation was seen to have been recovered,though not fully. On the other hand, the results obtained by
Hudacsek and Bransby [9] indicated that the rate ofdownlsope deformation reduced greatly during drying but it
did not show any significant recovery. The tests were carriedout under condition similar to that of Take and Bolton [7].
Field studies had also been carried out (e.g., Lim et al [10];Springman et al [11], Li et al [12]) with the focus mainly on
the mechanisms of rainfall infiltration and their influence on
slope stability in various ground conditions. Studies had
further shown that the slopes exhibit seasonally fluctuationand the downslope movements accumulates (Malet [13];Coe[14]; Malet [15]; van Asch [16]; van Asch [17];Glastonbury [18]; van Asch [19]). Cheuk et al [20] monitored
a high slope (height of 17m) having a slope angle of 25o in Tai
Lam Chung, Hong Kong. The geology comprises layer ofcolluvium (1-2m thick) overlying completely decomposed
volcaniclastic and tuffaceous sediments (approximately 30m
thick). As reported by Cheuk el al [20] :2. CASE ONE (Kwong and Leung [21,22]
Figs. 1 and 2 present the plan of the site and the geologicalfeature at a typical section. The bed rock is moderately strong
to strong yellowish brown to brownish grey, coarse ash tuff.
The rockhead is highest, about 40~45 mPD at the crest of theslope and dips in a southwest direction. Hence, the rockhead
is dipping parallel with that of the cut slope. The rock,depending on its depth below the surface, has been
decomposed to various stages and can be categorised into
three layers: colluvium layer, completely decomposed
volcanic (CDV) layer and highly decomposed volcanic(HDV) layer. The CDV soil is characterized by higher
5

The daily rainfall records in the same period are also given in
the figure. As expected, the groundwater table rose in the wet
season and dropped in the dry season. The largest fluctuationof groundwater level was about 9 m and it was recorded at
DH213.
In Fig. 4, the variations of slope movement at an inclinometer(I2) with time are plotted. One can easily note the strong
correlation between the slope movement and the groundwaterlevel. The slope moved in the downhill direction
progressively as the groundwater level rose during the rainyseasons. On the other hand, there were considerable reboundmovements during the dry seasons when the groundwaterlevel dropped. Fig. 5 depicts the variations of horizontal
downslope displacements with depth for the inclinometer.
Note that the corresponding records at the start of the dry
season (1 October) in 1992 were taken as the reference.
During the first dry period (October 1992 to April 1993), theinclinometer readings rebound. In particular, there was an
overall upslope movement for the upper portion of the soil.
The peak rebound displacement was 17 mm, at a depth of 3 m
underneath the ground surface.
During the first wet period (May 1993 to September 1993),
one can note that significant downhill displacements were
mobilized in the slope by the rainfall.
The measurements at other inclinometers showed quite
similar trend with larger displacement mobilized at shallow
depth [24].

percentage of soil particles having smaller grain size, whilst
relatively more coarse particles are contained in the HDV
soil.
One can also note that the upper part of the slope was formedby cutting into the natural ground and the lower part is a cribwall retaining compacted backfill. The crib wall is about 5 mhigh. A closer examination of the rockhead level in the areaaround shows that there is a localized depression of therockhead level.
As the slope was subjected to severe deformation in the formof surface cracks and local failures during the previous rainyseasons, sensors, including piezometers and inclinometers,
was installed to monitor the slope. Daily rainfall during theperiod was also monitored by raingauges.
3

L

BLOCK 2

TAI PING VILLAGEBLOCK 1

38.3

66.7

Cul

E

60

56

52

CH

.5
37

RO
AD

Cul
29.6

64

LO

28.2

68.7

Baske
tb
Court all

43.0

Cul

RO

PO

.8
66

AD

Cul

64
60

Pak Kong Water
Treatment plant

23.6

A

I4
DH213
56
56
I1

DH206
I2
I3

57.2

55

50.2
DH208

50

40

RO
AD

37.0

Filter Bed

Filter Bed

study area

29.6

Water Tank
Water Tank

AD

0

25

50 m

Ro
ad

RO
23.5

400

46
O LONGVILLAGE

DH209

A

Water
Tank

72
70

Water
Tank

300

DH213
45

40

200
DH206

DH208

35

Daily rainfall (mm)38.1

Groundwater level (mPD)

45.1

100

DH209
30

Fig. 1 Plan of the slope

2593-01-01

0
93-05-01

93-09-01

94-01-01

94-05-01

94-09-01

95-01-01

95-05-01

Date (yy-mm-dd)

Fig. 3 Time variation of groundwater levels at different
drillhole locations in relation to the daily rainfall records
In the subsequent dry periods (October 1993 to April 1994
and October 1994 to April 1995), the inclinometer readings
indicate that the slope bounced backwards after the preceding
wet periods (April to October in 1993 and 1994).
Fig 2 Typical geological cross section of the slope
2.1 Monitored results
Fig. 3 presents the variations of groundwater level measuredby piezometers between 15 December 1992 and 31 May 1995.
6

200

layers was modelled by a nonlinear hypoelastic-perfectly
plastic model with a non-associated flow rule. Readers canrefer to Zhou et al [24] for the details of the adopted
constitutive relationships.
Fig. 8 presents the variation of relative rebounddisplacements due to drying with depth predicted by thenumerical model. The corresponding field results are also
shown in the same figure for comparison. As mentioned in theprevious section, the horizontal displacement on 06
September 1994 (start date) was taken as the reference for thedetermination of the relative rebound movement.
On 01 November 1994, the simulation results show
considerable rebound movements in response to the changeof pore water pressure redistribution induced by a drop ofgroundwater level of about 2.5 m. The predictions give
similar distribution pattern along the depth direction as thecorresponding field measurements, with a slightover-prediction. Furthermore, both the predictions and fieldmeasurements illustrate that the affected range of soil layers
representing rebound feature greatly exceeds the thickness of
the unsaturated zone.
With the drying progressed from November 1994 to March
1995, both the field and numerical results (Fig. 8) present asignificant increase of rebound deformation. It has to be
pointed out that the change of the groundwater level in thisstage was about twice as large as that during the previous
stage. However, the predictions and field monitoring data
demonstrate that only about 60% increase in the incremental
rebound displacements is induced by the redistribution ofpore water pressure.

50

Horizontal displacement (mm)150

Upslope movement

45

100
40
50

35

Groundwater level (mPD)

Downslope movement

0
GWL at DH208-50
93-01-0130
93-05-01

93-09-01

94-01-01

94-05-01

94-09-01

95-01-01

95-05-01

Date (yy-mm-dd)

Fig. 4 Time variation horizontal displacements at a depth of
0.5m below the ground surface in relation to the change ofgroundwater level (I2)
Horizontal displacement (mm)
-50
0

0

50

100

150

200

5

Depth (m)

10

15

20

25

zeroline
01/12/1992
30/03/1993
28/09/1993
29/03/1994
06/09/1994
22/03/1995

30

Fig. 5 Distribution of horizontal downslope displacementwith respect to depth during the monitored period (Thereadings at 01/10/1992 are taken as the reference) (I2)
2.2 Numerical simulation
The finite element mesh for the simulation is shown in Fig. 6.
The four-node bilinear plane strain element was used in themodel. Each node of the soil elements consisted of threedegrees of freedom (one for pore water pressure and two fordisplacements). On the other hand, only deformation degrees
of freedom were considered in the elements for modelling the
cribwall, the rockfill, and the bedrock layer, as theredistribution of pore water within these layers should be
negligible. During the hydraulic analysis for a steady statesolution of the pore water pressure field, the groundwaterlevel distribution at the corresponding stage in the slope wasdefined in the numerical model (Fig. 7). With the calculated
pore water pressure distribution, a mechanical analysis was
conducted subsequently to determine the slope deformation
with the displacement boundary conditions given in Fig. 6.
For the cribwall and rockfill materials, as well as the bedrock
layer underneath the HDV layer, the deformation was smallenough during the monitored period. Hence, they weremodelled by a linear elastic model with the parameters can befound in Table 1 of Reference [21]. The behaviour of the soil

Fig. 6 Finite element model of Section A-A of the study slope

Fig 7 Groundwater level distributions

7

Bounce-back displacement (mm)
0

-25

-20

-15

-10

-5

Past record (Fig. 10) indicates that the slope was moving
slowly downhill and the occurrence of landslides was not
infrequent. The movement is believed to be occurring along
the weak interface between colluvium and completely
decomposed andesite or within the completely decomposed
andesite layer.

0-5Depth (m)

-10

-15

-20

-25

By Dry I Measurements
By Dry II
By Dry I
Predictions
By Dry II-30

Fig 8 Comparison of the predictions and the monitored
horizontal bounce-back movements (The results on
06/09/1994 are taken as the reference)
Fig. 10 Landslide History

It can also be observed that some discrepancy exists between
the numerical predictions and the field measurements, and it
could be attributed to the two-dimensional plane strain
simplification in the numerical model. It should also be notedthat as only limited piezometric data were obtained from field
monitoring (Fig. 6), differences may exist between theprescribed groundwater distributions in the numerical modeland the actual conditions.

3.1 Monitored results
To closely monitor the slope, “real-time” monitoring was
conducted.
The instrumentations include piezometers,
inclinometers, moisture probe, tensiometers and rain gauges.
The layout for the inclinometers is shown in Fig. 11 and thatfor other instruments can be found in Xu [23].

3. CASE TWO (Xu [23])The second case involves a relict old landslide resting on a
V-shaped valley. The typical section is shown in Fig. 9. It ismuch steeper in the upstream than the downstream (the slope
angle is approximately 20o). The top stratum is a thin layer ofcolluvium. The typical thickness of colluvium is from 1.3m to
5.7m; the average thickness is approximately 3m. Under thecolluvium layer is the highly sheared and altered completely
decomposed andesite which has a depth of around 20m. The
bottom stratum is bedrock which consists of moderately
decomposed Andesite and slightly decomposed andesite.

BH11

Fig. 11 Layout Plan of In-place Inclinometers
Fig. 12 shows the variation of the groundwater table at
mid-slope (BH11) from 2008-2009. At the beginning of thewet season, the groundwater level was about 136mPD. One
can note the groundwater rose rapidly in response to therainfall and it could have nearly reached the ground surface
after prolong period of heavy rain during the wet season. On
the other hand, the groundwater table dropped in the dry
season gradually and it could have almost dropped back to thelevel at the beginning of the wet season. The groundwatertable along the slope is depicted in Fig. 13.

Fig. 9 Typical Geological Section

8

Ground Level(+142.23mPD)
120
100
80
60
40
20

Hourly Rolling Rainfall (mm)TSF-BH11(+127.83mPD)

02008-6-1
2008-6-3
2008-6-5
2008-6-7
2008-6-9
2008-6-11
2008-6-13
2008-6-15
2008-6-17
2008-6-19
2008-6-21
2008-6-23
2008-6-25
2008-6-27
2008-6-29
2008-7-1
2008-7-3
2008-7-5
2008-7-7
2008-7-9
2008-7-11
2008-7-13
2008-7-15
2008-7-17
2008-7-19
2008-7-21
2008-7-23
2008-7-25
2008-7-27
2008-7-29
2008-7-31

Level (+mPD)

Hourly Rolling Rainfall
143
142.5
142
141.5
141
140.5
140
139.5
139
138.5
138

Fig.15 Groundwater table (IPI3)

Fig. 12 Groundwater Levels Monitored at BH11

Fig.13 Groundwater distribution
The deformations at IPI3 (Figs. 14 and 15) also indicate thatthe slope deformation is closely correlated to the rainfall andthe groundwater level. The slope moved downhill during thewet season, though the magnitudes depended on the depth
below the slope surface : most of the movement was restricted
to the top 3-4m. The movement also indicates that the slope
rebound after rain and the uphill movement continued during
the dry seasons. However, the movement was not fully
recovered. Similar movement pattern was observed in otherlocations, though the magnitude of the downhill and uphill
movements could be different. The general direction of the
movement is shown in Fig. 16 and one can note that the slope
is moving downhill along the valley.

TSF-IPI3-A3

TSF-IPI3-A4

3.2 Numerical simulation
A 2-D finite element mesh was also built up according to thegeometry of typical section of the slope for numericalanalyses. The length of the model was 450 m, and the heightof the model was 150m; there were typically three layers:
colluvium (including talus at slope head), completely
decomposed andesite (CDA) and bedrock. The slope wasmodelled using a finite element mesh consisting of 3122
nodes and 2784 plane-strain four-node bilinear elements.
Again, each node for the soils has three dofs (degree of
freedom): one was for pore water pressure and the other two
were for displacements. The controlling sections and points
are shown in Fig. 17. In the analyses, various boundary
conditions and rainfall patterns were considered andinterested readers can refere to Xu [23] for the full details.
The present paper will focus on the movement and porepressure at Section E.

TSF-IPI3-A5
150
140
130
120
110
100
90
80
70
60
50
40
30
20
10
0

Hourly Rolling Rainfall (mm)Cumulative Deflection (mm)Hourly Rolling Rainfall
110
100
90
80
70
60
50
40
30
20
10
0
-10
-20

Fig.16 Schematic Deformation Mode

Fig.14 Downslope Cumulative Deformation Monitored(IPI3)

9

mPD
+195

+155

BH11

+150
+145

IPI1

+140
+135

E

BH13

AGD2
IPI6

+130
+125
+120

IPI3

AGD3

+115
+110

IPI4

+105

10m

BH14

+100
+95
+90

Fig.17 Monitored Groundwater Distribution Illustration at
Typical Section S1-S1

140

300
250

100

0
60

50
0
170

June 1

175

180

185

Day

190

195

170

175

180

185

Day

190

195

200
June 30[1] Li AG, Yue ZQ, Tham LG, Lee CF, Law KT,“Field-monitored variations of soil moisture andmatric suction in a saprolite slope,” Canadian
Geotechnical Journal, vol 42(1), 2005, pp 13-26.
[2] Zhan TLT, Ng CWW, Fredlund DG, “Field study ofrainfall infiltration into a grassed unsaturated
expansive soil slope,” Canadian Geotechnical
Journal, vol 44(4), 2007, pp 392-408.
[3] Olivares L, Damiano E, “Postfailure mechanics oflandslides: Laboratory investigation of flowslides inpyroclastic soils,” Journal of Geotechnical andGeoenvironmental Engineering, vol 133(1), 2007,
pp 51-62.
[4] Melinda F, Rahardjo H, Han KK, Leong EC, “ Shearstrength of compacted soil under infiltration
condition,” Journal of Geotechnical andGeoenvironmental Engineering, vol 130(8), 2004,
pp 807-817.
[5] Farooq K, Orense R, Towhata I, “Response of
unsaturated sandy soils under constant shear stressdrained condition,” Soils and Foundations, vol44(2), 2004, pp 1-13.
[6] Collins BD, Znidarcic D, “Stability analyses ofrainfall induced landslides,” Journal of Geotechnicaland Geoenvironmental Engineering, vol 130(4),2004, pp 362-372.
[7] Zhang LL, Fredlund DG, Zhang LM, Tang WH,“Numerical study of soil conditions under which

0

100

50

6. REFERENCES

Daily Rainfall Intensity (mm)

45
30
15

200
150

-15

Pore Water Pressure Change (kPa)

250

150

5. ACKNOWLEDGMENT
The authors would like to acknowledge the contribution ofProfessor Zhou Yuande of Tsinghua University, China and
Dr AKL Kwong of MTRC, HK.

450

300

200

Fig.19 Comparasion between Simulated and Monitored
Horizontal Displacement of Case T3 at Section E

500

350

350

June 1

4. CONCLUSIONSThis paper presents a study on the rebound behaviour
observed in two slopes. Field measurements of deformation,
groundwater level as well as the rainfall intensity were
presented.and the monitoring data indicated that considerable
bound-back deformation in the uphill direction were
mobilized in the slope during the dry seasons. Simulations,based on two-dimensional finite element model, were carriedout and typical wetting-drying processes had been analyzed
to investigate the mechanism of the drying induced reboundmovements. Through comparison with the corresponding
field measurements, it is found that the volume contractive
behaviour, as a result of the suction increase within theunsaturated soil zones and the pore pressure reduction in thesaturated soil zones is the main factor contributing to suchrebound movements.

400

400

0

The comparisons of the simulated and measured pore waterpressure and horizontal displacement are shown in Figs. 18
and 19, respectively. Though there are discrepancies in themagnitude for both parameters, the simulation can provide
quite a reasonable prediction of their variation qualitatively.
It appears that some fine tuning of the input parameters canimprove the prediction, however, matching of the results is
not the objective of the present study.

Daily Rainfall Intensity - June of 2008
Simulated Pore Water Change at section E
Monitored Pore Water Change at section E(AGD3)

450

Daily Rainfall Intensity (mm)D

Dry season120

At May 30,2008

+160

500
Daily Rainfall Intensity - June of 2008
Simulated Horizontal Displacement at the Surface of Section E
MonitoredHorizontal Displacement at the Surface of Section E (IPI6)

100

+165

C

80

+170

60

B

+175 The highest W.T

40

+180

Cumulative Horizontal Displacement (mm)

A

+185

20

+190

200
June 30

Fig.18 Comparison between Simulated and Monitored Pore
Water Change of Case T3 at Section E
10matric suction can be maintained,” Canadian
Geotechnical Journal, vol 41(4), 2004, pp 569-582.
[8] Take WA, Bolton MD, “Identification of seasonal
slope behaviour mechanisms from centrifuge casestudies,” Advances in Geotechnical Engineering:
The Skempton Conference, London, Thomas
Telford, vol. 2, vol 2004, pp 992-1004.
[9] Hudacsek P, Bransby MF, “Centrifuge modelling ofembankments subject to seaonsal moisturechanges,” Proceedings of the First ISSMGEInternational Conference on Transportation
Geotechnics, Nottingham, UK, September, 2008.
[10] Lim TT, Rahardjo H, Chang MF, Fredlund DG,“Effect of rainfall on matric suctions in a residual
soil slope,” Canadian Geotechnical Journal, vol33(4), 1996, pp 618-628.
[11] Springman SM, Jommi C, Teysseire P, “ Instabilities
on moraine slopes induced by loss of suction: A casehistory,” Geotechnique, vol 53(1), 2003, pp 3-10.
[12] Li J, Tham LG, Junaideen SM, Yue Zq, Lee CF, “Loose fill slope stabilization with soil nails:
full-scale test,” Journal of Geotechnical andGeoenvironmental Engineering, vol 134(3), 2008,
pp 277-288.
[13] Malet JP, Maquaire O, Calais E, “The use of global
positioning system techniques for the continuous
monitoring of landslides: application to Super-Sauze
earthflow (Alpes-de-Haute-Provence, France),”
Geomorphology, vol. 43, 2002, pp. 33-54.
[14] Coe JA, Ellis WL, Godt JW, Savage WZ, Savage JE,
Michael JA, Kiblet JD, Powers PS, Lidke DJ,
Debray S, “Seasonal movement of the Slumgullion
landslide determined from global positioning systemsurveys and field instrumentation, July 1998-March
2002,” Engineering Geology, vol. 68, 2003, pp.
67-101.
[15] Malet JP, Laigle D, Remaitre A, Maquarie O,“Triggering conditions and mobility of debris flows
associated to complex earthflows,” Geomorphology,
vol. 66, 2005, pp. 215-235.
[16] van Asch TWJ, “Modelling the hysteresis in thevelocity pattern of slow-moving earth flows: the roleof excess pore pressure,” Earth Surface Professessand Landform, vol. 30, 2005, pp 403-411.
[17] van Asch Th W J, Van Beek L P H, Bogaard,“Problems in predicting the mobility ofslow-moving landslides,” Engineering Geology, vol91, 2007, pp 46–55.
[18] Glastonbury J, Fell R, “Geotechnical characteristics
of large slow, very slow, and extremely slow
landslides,” Canadian Geotechnical J, vol 45, 2008,
pp 984-1005.
[19] van Asch Th W J, Malet J P, Bogaard, “The effect ofgroundwater fluctuations on the velocity pattern ofslow-moving landslides,” Nat. Hazards and Earth
System Sciences, vol 9, 2009, pp 739–749.

[20] Cheuk J, Ng A, Endicott J, Ho K, “Progressive slope
movement due to seasonal wetting and drying,” The
State-of-the-art Technology and Experience on
Geotechnical Engineering in Malaysia and Hong
Kong Wednesday, 25 February 2009, pp 105-114.
[21] Kwong AKL, Leung RHT, “Back analyzing the
failure mechanism of a slope experienced large
movements from 1986 to 1997,” International
Conference on Slope Engineering, Hong Kong,edited by CF Lee and LG Tham, 2003, pp 796-802.
[22] Kwong AKL, Leung RHT, “Constructional aspectsof remedial works for slope above Pak Kong water
treatment works,” International Conference on
Slope Engineering, Hong Kong, edited by CF Lee
and LG Tham, 2003, pp 803-809.
[23] Xu K, “Study on a Slow Moving Slope in Hong
Kong,” PhD Thesis, Department of Civil
Engineering, The University of Hong Kong, 2011.
[24] Zhou Y D, Tham L G, Kwong A K L, Tang X W,
“Mechanism of drying induced rebound movements
in a soil slope in Sai Kung, Hong Kong,”
Engineering Geology, vol 116(1-2), 86-94, 2010.

11

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Bearing Capacity and Settlement of Footings in Unsaturated Sands
Sai K. Vanapalli1 and Fathi M. O. Mohamed2
Department of Civil Engineering, University of Ottawa, K1N 6N5, Canada

1,2

ABSTRACTThe focus of the study presented in this paper is towards understanding the influence of three parameters; namely, (i)matric suction, (ii) overburden stress, and (iii) dilation, on the bearing capacity and settlement behaviour of surface andembedded footings in saturated and unsaturated sands. The results show that the bearing capacity and settlement
behaviour of unsaturated sands are significantly influenced by all the three parameters. In addition, comparisons areprovided between the measured and estimated bearing capacity and settlement using the proposed modified Terzaghi’s
equation and modified Schmertmann’s CPT-based method, respectively. There is a good comparison between themeasured and estimated values for both the laboratory and field tests using the modified methods.
Keywords: Bearing capacity; settlement; matric suction; sand; PLT; CPT.

1.

INTRODUCTION

The two key properties required in the design of shallow
foundations are the bearing capacity (i.e., q u ) and settlement(i.e., ) behaviour of soils. Structures such as silos, antenna
towers, bridges, power plants, and housing subdivisions canbe constructed on shallow foundations (e.g., spread footingsnear the ground surface) in sandy soils assuring a safe andeconomical design. The shallow footings are typicallydesigned to transfer the loads safely from the superstructure
to the supporting soil such that the settlements are in
acceptable limits as per the design and construction codes.
The bearing capacity of shallow foundations is estimatedusing the approaches originally presented by [1] and [2]assuming the soil is in a state of saturated condition. Typically,shallow foundations are placed above the ground water table
and the variation of stress with respect to depth associatedwith the loads from the superstructure is distributed throughthe substructure (i.e., shallow foundations) predominantly inthe vadose zone (i.e., the zone in which soil is in a state of
unsaturated condition). This is true in many regions andespecially for semi-arid and arid regions. A framework forinterpreting the bearing capacity and settlement behaviour
from experimental and modelling studies for unsaturated
sands is recently evolving ([3], [4], [5], [6], and [7]).
Comprehensive data for interpreting the bearing capacity andsettlement behaviour of footings in unsaturated sands taking
account of influence of the capillary stresses (i.e., matricsuction), overburden stress (i.e., confinement) and dilation is
however are not available in the literature. Due to thesereasons, an extensive experimental program is undertaken to
study the bearing capacity and settlement behaviour of sandy
soil from model footings in specially designed equipment.

12

In addition, comparisons are provided between the measured
and estimated values of the bearing capacity and settlement
behaviour of model footings respectively using theapproaches provided by investigators [3] and [7] for sands
modifying the original contributions of Terzaghi’s equation[1] and Schmertmann’s equation [8], respectively. The study
shows that there is a good comparison between the measured
and estimated values of the bearing capacity and settlement
behaviour using the proposed modified equations.
2.

BACKGROUND

2.1 Bearing Capacity of Soils
Terzaghi and others [1], [2], [9] studies were directed towards
understanding the bearing capacity of shallow foundations insaturated or dry conditions using conventional soil
mechanics. However, soils are typically found in a state of
unsaturated condition in semi-arid and arid regions.
Therefore, estimation of the bearing capacity of shallow
foundations using conventional soil mechanics for theseregions may underestimate the bearing capacity values andlead to conservative and costly foundation designs.
Several researchers carried out investigations to study thebearing capacity of unsaturated soils ([10], [11], [12], [13],[14], [3], [15]). Investigators [16] designed special equipment
and conducted studies to understand the bearing capacity ofsurface model footing in a sandy soil. These studies haveshown that the matric suction values in the range of 2 to 6 kPa
contribute to 5 to 7 times higher than the bearing capacity
values in comparison to saturated condition. A framework
was provided by [3] to predict the variation of bearing
capacity of a soil with respect to matric suction using thesaturated shear strength parameters (c' and ') and theSoil-Water Characteristic Curve (SWCC).

dilatancy behaviour of sands ([9], [19], and [20]). The
dilatancy angle,
 is always
angle, ' based on the studies by [21]. The dilation
behaviour of sand can be attributed to the soil particles rollingon top of each other without crushing during the shearing
stage. Experiments (using steel shots that do not break down
during shearing) conducted by researcher [22] have shownthat an increase in confinement leads to a decrease in the
dilatancy angle
 . Modilation of sand decreases with an increase in the effective
overburden stress [23]. Therefore, in the analysis of surface
model footing results of the present study, bearing capacity
and settlement behaviour of shallow footings were interpreted
taking into account of the influence of dilation angle,
 on theeffective friction angle, ' of the sand.

2.2 Settlement of Shallow Footings
The shallow footings are typically designed in sandy soils
such that the settlement is less than 25 mm. In addition,
applied loads from the superstructure need to be safely carried
to the soil below the footing with a factor of safety
recommended by the design and construction codes. Elastic
or immediate settlements in sandy soils are assumed to occur
instantaneously when static loads are applied. Several
empirical equations are proposed in the literature that can beused in the estimation of the settlement of footings in sands
based on cone penetration tests (CPT) results ([17], [9], and[8]). The presently available methods in the literature
overestimate the settlements leading to an overly
conservative footing design ([18], and [7]). This can be
attributed to ignoring the influence of matric suction below
the foundations. In many scenarios such an assumption is not
valid, particularly where the soils are typically in a state of
unsaturated condition.
Simple relationships are proposed by [7] modifying theSchmertmann’s method [8] that is conventionally used inpractice for settlement estimations from the CPT results. The
modified method was successfully used in the estimation of
the settlement behaviour of model footing tests and full-scalefootings tested in-situ under both saturated and unsaturated
conditions in sandy soils.
The focus of the present study is to understand the influenceof the capillary stresses (i.e., matric suction), overburden
stress (i.e., confinement) and dilation on the variation of the
bearing capacity and settlement behaviour of both surface andembedded footings in unsaturated sand. In the present study,
the same sand used by investigators [3] is tested.
3.

4.

Fig. 1 shows the details of the modified University of Ottawa
Bearing Capacity Equipment (modified UOBCE). Schematicof a sectional view of the equipment is shown in Fig. 2. The
modified UOBCE is designed to determine the variation of
bearing capacity and settlement of sands with respect tomatric suction using model footings which are interpreted
similar to plate load tests (i.e., PLTs). In the remainder of thepaper, model footing tests are referred to as model PLTs forbrevity. The equipment setup consists of a rigid-steel frame
made of rectangular section pipes with thickness of 6 mm anda box of 1500 mm (length) × 1200 mm (width) × 1060 mm(depth). The test box can hold up to 3 tons of soil and thecapacity of the loading machine (i.e., Model 244 HydraulicActuator with stroke length of 250 mm) is 28.5 kN. The
model PLTs were performed using different strain rates of 1.2
mm/min and 2.5 mm/min. The results suggest that the load
carrying capacity of the sand is not influenced by the twodifferent strain rates used in the present study.
The equipment used in the present study (see Fig. 1) in termsof test box size and its loading capacity is twice in comparison
to the UOBCE used by [16]. The equipment in the present
study has special provisions to achieve different degrees ofsaturation conditions below the model footings similar to theoriginal UOBCE. The variation of matric suction with respect
to depth in the unsaturated zone of the test box can bemeasured using commercial Tensiometers.

THE TESTED SOIL

3.1 Soil Properties
Table 1 summarizes the properties of the sand used in thisstudy. The soil can be classified according to the USCS as
poorly graded coarse-grained sand.
Table 1. Properties of the tested soil
Soil properties
Coefficient of uniformity, C u
Coefficient of curvature, C c
Specific gravity, G s
Average dry unit weight, d , kN/m³
Min. dry unit weight, d (min) , kN/m³
Max. dry unit weight, d (max) , kN/m³
Average relative density, %Optimum moisture content, OMC, %Void ratio, e (after compaction)
Effective cohesion, c' , kN/m2
Effective friction angle, ' (o)EQUIPMENT AND METHODOLOGY

Value
1.83
1.23
2.65
16.02
14.23
17.25
65.0
14.6
0.62 – 0.64
0.6
35.3

5.

LABORATORY PLT AND CPT TESTS

Several tests were conducted to determine the bearing
capacity of the sand with different values of matric suction
using surface PLTs (i.e., model footing depth, D f = 0 mm),embedded PLTs (i.e., D f = 150 mm) and CPTs (i.e., cone
penetration tests). A minimum of three tests were conducted
and average values are reported in this paper.
5.1 Surface Plate Load Tests
Model PLTs of 150 mm × 150 mm (i.e., surface footings)
were conducted by researchers [16] using the UOBCE (seesectional view in Fig. 2) which consists of a rigid-steel tank of
900 mm (length) × 900 mm (width) × 750 mm (depth).
Applied stress versus settlement relationships for surface

Dilation in Sandy Soils
The effective cohesion, c' and the angle of internal friction,
' were measured from the direct shear test (i.e., CD test)
results. Several studies suggest the overburden effective
stress (i.e., confinement) and soil density influence the13model footing of 150 mm × 150 mm from that study aresummarized in Fig. 4.
Tensiometer
Compacted sand

MTS Actuator
Load cellFig. 2. Schematic to illustrate the test setup and the procedureused for estimating the average matric suction of 6 kPa within
the stress bulb (SB) zone of the surface footing using UOBCEDrainage valve

Model footing
Piezometer

Fig. 1. Modified University of Ottawa bearing capacity
equipment (modified UOBCE)
5.2 Embedded Plate Load Tests
The model footings embedded in the modified UOBCE test
box are analyzed considering the influence of average matricsuction value in the proximity of the stress bulb zone which isequal to depth 1.5B. Fig. 3 provides details of the procedureused in the estimation the average matric suction value. The
depth 1.5B considered is the zone in which stresses are
predominant due to the loading of shallow square footings
with D f /B ≤ 1.0 ([24], [3], and [5]).

Fig. 3. Schematic to illustrate the test setup and the procedureused for estimating the average matric suction of 6 kPa within
the stress bulb (SB) zone of the embedded footing using
modified UOBCEIn this series of tests, the model footing of 150 mm × 150 mm
size is placed at a depth of 150 mm below the soil surface toinvestigate the effect of the overburden stress. The tests were14

conducted with different average matric suction values below
the footing (i.e., 0 kPa, 2 kPa and 6 kPa).
Equilibrium conditions with respect to matric suction in thetest box were typically achieved in a period of 48 hrs in thetest box shown in Fig. 3. Table 2 summarizes a typical set ofresults in which the average matric suction in the vicinity of
the footing base is 6 kPa (see Fig. 3).

5.3 Cone Penetration Tests
Researchers [25] conducted several CPTs in a laboratory
Compacted sand

LVDT

Tensiometer

1400
Footing depth = 150 mm
Matric suction = 6 kPa
Footing depth = 150 mm
Matric suction = 2 kPa

2Applied stress, q (kN/m )

1200
1000
800

Footing depth = 0 mm
Matric suction = 6 kPa

600

Footing depth = 0 mm
Matric suction = 2 kPa

400

Footing depth = 150 mm
Matric suction = 0 kPa

200Footing depth = 0 mm
Matric suction = 0 kPa

0
0

10

20

30

Cone penetrometer

40

Settlement, δ (mm)Water supply

Fig. 5. Top view of the test tank showing the penetrometer(CPT) using the UOBCE

Fig. 4. Relationship between the applied stress versus
settlement behaviour of surface and embedded model footing
tests (PLTs) of 150 mm × 150 mm using the UOBCE and themodified UOBCE respectively
Table 2. Typical data from the test box for AVR matricsuction of 6 kPa in the stress bulb zone (i.e., 1.5B)
D*
(mm)

t

(kN/m3)d(kN/m3)

w
(%)

S
(%)

12
18.16
16.20
12.10
53
150
19.00
16.24
17.00
75
355
19.20
16.13
19.00
82
500
19.50
16.12
21.00
91
700
19.74
16.03
23.11
98
800
19.75
15.95
23.81 100*
Depth of a Tensiometer from the soil surface(u a u w ) AVR
(kPa)
8.0
7.0
5.0
2.0
1.0
0.0

The measured water content and matric suction values fromthe test box of the modified UOBCE are similar to thecorresponding water content and matric suction values inboth measured and predicted SWCC of the tested sand. Thisobservation provides credence to measurements using theTensiometers. The air-entry value for the sand was found tobe between 2.5 kPa and 3 kPa. Details of the SWCC for thetested sand are available in an investigation conducted byresearchers in [16].
The measured bearing capacity of the compacted unsaturated
sand for both surface and embedded footings were in therange of 5 to 7 times higher than the saturated bearing
capacity values.

Fig. 6. Schematic to illustrate the test setup and the procedureused for estimating the average matric suction of 6 kPa within
the influence zone (IZ) using the model penetrometer (CPT)using the UOBCE
15

environment in a compacted sand (D r = 65 %) in the UOBCE
under both saturated and unsaturated conditions (i.e., 0 kPa, 1
kPa, 2 kPa and 6 kPa). The setup, experimental results andanalyses of the variation of cone resistance, q c withpenetration depth were presented in [25]. Fig. 5 and Fig. 6
show details of the sectional view of the test box used to carryout the CPTs respectively. The measured settlement results ofthe studies described in section 8.2 are used to check thevalidity of the proposed modified Schmertmann’s
relationships based on the CPTs results in a later section toestimate the settlement of shallow footings in sand under bothsaturated and unsaturated conditions.
6.

into account of the influence of overburden stress and theshape factors.
qu =[c '+ (ua − uw )b (tan φ '− Sψ BC tan φ ')+(ua − uw ) AVR Sψ BC tan φ '] N c ζ c Fc
+γ D f N q ζ q Fq + 0.5 γ B Nγ ζ γ Fγ

where:
F c , F q , F = depth factor
q= shape factor

BEARING CAPACITY OF UNSATURATEDSOILSTerzaghi [1] suggested Eq. (1) to estimate the ultimate bearing
capacity, q u of saturated soils assuming general shear failure
conditions:qu =c ' N c + γ D f N q + 0.5γ B Nγ

(1)

= ultimate bearing capacity, kN/m2= effective cohesion, kPa= unit weight, kN/m3
Df= footing base level, mB= footing width, m
N c , N q , N  = bearing capacity factors which are function ofeffective friction angle, '.
A semi-empirical equation (i.e., Eq. (2)) was suggested by [3]to predict the variation of bearing capacity with respect tomatric suction for surface footings on unsaturated soils using
the effective shear strength parameters (i.e. c' and ' ) andthe SWCC as below:+ (ua − uw ) AVR Sψ tan φ '] N c ζ c + 0.5γ B Nγ ζ γ

The bearing capacity fitting parameter, ψ BC along with theeffective shear strength properties (c' and ' ) and the
SWCC are required for predicting the variation of bearing
capacity with respect to matric suction assuming drained
loading conditions. The bearing capacity fitting parameter,
ψ BC can be estimated from relationship provided byresearchers [3] in Eq. (4) given below:
ψ BC =
1.0 + 0.34 ( I p ) − 0.0031 ( I p 2 )where:quc'

qu =[c '+ (ua − uw )b (tan φ '− Sψ BC tan φ ')

(3)

(2)

where:(u a - u w ) b = air entry value from SWCC, kPa(u a - u w ) AVR = average matric suction value, kPa (Fig. 2; Fig.
3)'= effective friction angle, ºS= degree of saturation, %BC= bearing capacity fitting parameterc , 
= shape factors(4)

Several investigators provided bearing capacity factors forcohesion, N c ; surcharge, N q and unit weight, N  ([1], [2],[26], and [26]). The values for bearing capacity factors of N c
and N q provided by various investigators are approximately
the same. For this reason, the bearing capacity factors, N c andN q originally proposed by Terzaghi [1] were used in theanalysis. The N values suggested by [26] have been more
widely used in recent years. For this reason, the bearing
capacity factor, N  values proposed by [26] are used in thisstudy.
7.

COMPARISON BETWEEN MEASURED AND
PREDICTED BEARING CAPACITY

7.1 Measured and Predicted B.C. for Surface PLTs
The bearing capacity values of surface model footings of 150
mm × 150 mm were measured using the UOBCE [16] underboth saturated and unsaturated conditions. Both the bearing
capacity under saturated and unsaturated conditions wasinterpreted taking into account of influence on the dilatancy
angle,
 for sand.
Table 3. B.C. factors, shape factors and depth factors used in
the analysis for the surface PLT
Effective friction angle, ' = 35.3º
Estimated dilatancy angle,  = 3.53º
Modified friction angle, ' m = (' + ) ≈ 39º
B.C. Factors1
Shape Factors2
Depth Factors3
N c N q N  ζ c q  F c
Fq
F
86 70
95
1.8 1.8 0.6 1.0
1.0
1.0
1
from [1]; 2 from [36]; 3 from [27]There is a smooth transition between the bearing capacity
equation proposed by [3] for unsaturated soils and theconventional Terzaghi’s bearing capacity equation forsaturated soils. In other words, the equation (i.e., Eq. (2)proposed by [3]) will be the same as Terzaghi’s bearing
capacity equation when the matric suction value is set equal
to zero.
The general form of equation to estimate the bearing capacityof unsaturated soils is shown in Eq. (3). This equation takes

The dilatancy angle,  was not measured in this study but
was approximated for typical sand based on the information
reported in the literature. The dilatancy angle,
 valueassumed to be equal to 10 % of effective friction angle, ' of16

35.3 is equal to 3.53º following [28] (i.e., DS 415. 1984). In
other words, the modified friction angle, 'm is 38.53º orapproximately 39º. Summary of the values of the bearing
capacity factors, shape factors and depth factors for thesurface model footings with a modified friction angle of 'm= 39° are presented in Table 3.

Table 4. B.C. factors, shape factors and depth factors used in
the analysis for the embedded PLT
Effective friction angle, ' = 35.3º
Estimated dilatancy angle,
 = 0ºB.C. Factors1
Shape Factors2
Depth Factors3
N c N q N  c q  F c
Fq
F
58 41
45
1.7 1.7 0.6 1.4
1.2
1.0
1
from [1]; 2 from [36]; 3 from [27]Residualzone

Transition
zone

Saturation
zone

1000

Measured and predicted B.C., qu (kN/m2)

2
Measured and predicted B.C., qu (kN/m )

1200

800
600
400
Measured bearing capacity for surface PLT

200

Predicted bearing capacityfor surface PLT

0

0

2

4

6

8

10

12

Matric suction, (ua - uw) kPa

Fig. 7. Measured and predicted B.C. for surface modelfooting (PLT) of 150 mm × 150 mm using the UOBCE

1400
Saturation
zone

1200

Transitionzone

Residual
zone

1000
800
600
Measured bearing capacity for embedded PLT

400

Predicted bearing capacity for embedded PLT

200
0
0

2

4

6

8

10

12

Matric suction (ua- uw) (kPa)

The same approach has been extended for analyzing surface
Fig. 8. Measured and predicted B.C. for embedded model
model footing results of another three sands tested byfooting (PLT) of 150 mm × 150 mm using the modified
researcher [11]. Similar to the test results of sand used in the
UOBCEpresent study, good comparison was observed between themeasured and predicted bearing capacity values for theseIn other words, the bearing capacity factors, shape factors andthree sands considering a dilatancy angle value equal to 10 %depth factors were obtained using the effective friction angle,of effective friction angle. Summary of these discussions are' = 35.3º (see Table 4). There is a good comparison
available in [29]. More recently, [5] have undertaken
between the measured and estimated bearing capacity values
numerical modeling studies to predict the variation of bearing
for interpreting the embedded model footing results without
capacity with respect to matric suction. These modelingtaking account of the influence of dilatancy angle,
 asresults also show that reasonably good comparisons wereshown in Fig. 8. Such a behavior can be attributed to thepossible between the measured and modeled bearing capacity
influence of the confinement with a depth which is equal tovalues using a dilatancy angle,
 which is equal to 10 % of
the width, B of the foundation (D f = 150 mm; as illustrated in
effective friction angle,  'as shown in Fig. 7.
Fig. 4). These results are also consistent with the studies ofseveral investigators who have shown that the influence ofdilation in the sand decreases with an increase in overburden
7.2 Measured and Predicted B.C. for Embedded PLTs
The bearing capacity of 150 mm × 150 mm embedded model effective stress or confinement ([19], [23], and [30]).
footing (sandy soils in both saturated and unsaturated
conditions) was measured using the modified UOBCE (see 8. CORRELATIONS BETWEEN THE CONERESISTANCE AND THE SETTLEMENT OF
Fig. 1). The model footing of 150 mm × 150 mm is located at
FOOTINGS IN SAND
a depth, D f of 150 mm below the soil surface simulating an
overburden stress which also acts as a confinement all around 8.1 Proposed Techniquethe footing.
The modulus of elasticity, E s typically increases with depth inEquation (3) is used in the interpretation of the bearing
sandy soils and the stresses associated with the applied load
capacity results of embedded footings in saturated anddecrease with an increase in depth. In other words, settlement
unsaturated sandy soils taking account of the influence of thewill be less in deeper layers in comparison to shallow layers.
overburden stress and the shape factors. However, theSchmertmann [8] suggested equation (i.e., Eq. (5)) by
influence of dilatancy angle,
 was not considered.
extending this philosophy for the estimation of footing
17

settlement in sands using average cone penetration resistance,q ci over a depth of 2B from the bottom of the footing.
2B δ = C1 C 2 (q a − σ ' z , d )∑0I ∆  zi zi 
 E

s 


Proposed technique for calculating Es(5)where:
C1 = 1 − 0.5 [σ ' z , d /(q a − σ ' z , d )] ; C 2 = 1 − 0.21 log [t / 0.1]

Saturated sandIf (ua - uw) = 0 kPa

Unsaturated sandIf (ua - uw) > 0 kPa

 = settlement, C 1 = depth factor, C 2 = time factor, q a =footing pressure; ’z,d = vertical effective stress at footingbase level; E s = f × q ci (i.e., elastic modulus of soil); I zi =influence factor, B = footing width; q ci = resistance of eachlayer; f = coefficient; zi = thickness of each layer; and t =time.

If Dr < 50 %f2 = 1.7 × ((Dr/100)2 + 3.75)

This method is widely used in geotechnical engineering
practice. One of the key limitations of this method is that it
does not take into account the influence of capillary stress or
matric suction and is used for sands both in saturated andunsaturated conditions.
Investigators [7] suggested two empirical relationships thatcan be used in the Schmertmann’s equation [8] to estimate thesettlements in sands. These empirical relations are useful inestimating the modulus of elasticity, E s of sands in saturated
and unsaturated conditions. The relationships are proposed
based on the analysis from the PLT and CPT results.
Equation (6) is suggested to estimate the modulus ofelasticity, E s for saturated sands (i.e., (u a - u w ) = 0 kPa) as
given below:
E s ( sat ) = f 1 × q c sat

f2 = 1.2 × ((Dr/100)2 + 3.75)

Es (unsat) = f2 × qc unsat

f1 = 1.5 × ((Dr/100)2 + 3)

Es (unsat)

Es (sat) = f1 × qc sat(6)
Es (sat)where: E s (sat) = modulus of elasticity for saturated
homogenous sand, f 1 = 1.5 × ((D r /100)2 + 3) (i.e. f 1 is acorrelation factor and D r is the relative density in % ), q c (sat) =average cone resistance under saturated sands condition (e.g.
within an influence zone, IZ equal to 1.5B from the footingbase level) and B = footing width.
Equation (7) is suggested to estimate the modulus ofelasticity, E s for unsaturated sands (i.e., (u a - u w ) > 0 kPa):
E s ( unsat ) = f 2 × q c unsat

If Dr ≥ 50 %Apply in Eq. (5)Fig. 9. Flow-chart to illustrate the proposed technique used
for calculating the modulus of elasticity, E s (sat) and E s (unsat)
Fig. 10 and Fig. 11 provide comparisons between theestimated and measured settlement values for both surface
and embedded respectively of 150 mm × 150 mm model
PLTs in the tested sand with different average matric suction
values (i.e., 0 kPa, 2 kPa and 6 kPa) using the proposed
relationships into the Schmertmann’s equation [8]. The
footing settlements decrease with an increase in the matricsuction and the overburden stress.

(7)

where: E s (unsat) = modulus of elasticity for unsaturated
homogenous sands, f 2 = 1.2 × ((D r /100)2 + 3.75) for sandswith D r < 50 % or f 2 = 1.7 × ((D r /100)2 + 3.75) for sands withD r ≥ 50%, (i.e. f 2 is a correlation factor and D r is the relative
density in %), q c(unsat) = average cone resistance underunsaturated sands conditions (e.g., within influence zone, IZ
equal to 1.5B from the footing base level) and B = footing
width.
Fig. 9 summarizes various steps detailed earlier in the form of
a flow chart. The modulus of elasticity, E s from Eq. (6) or Eq.
(7) can be substituted into Schmertmann’s equation (i.e., Eq.
(5)) [8] to estimate the immediate settlement.
18

unsaturated sand from in-situ presented in [35]. It can be seenthat the proposed technique provides reasonable settlement
estimations in comparison to the measured values. This canbe attributed to the use of different correlations (i.e., f 1 and f 2 )between the cone resistance, q c and modulus of elasticity, E s
in the new relationships which are better in representing thesettlement behaviour of sands. Furthermore, the selection ofthese correlations are not only functions of the CPTs results
but also are based on the soil condition (e.g., saturated or
unsaturated) and the relative density, D r .

Applied stress, qu (kN/m2)

1000
800
600

Measured δ for 6 kPa

400
Estimated δ for 6 kPa

200

1000
0

5

10

15

20

25

Applied stress, qu (kN/m2)

030

Measured or estimated settlement, δ (mm)Fig. 10. Comparison between the measured and estimatedsettlements using the modified Schmertmann’s equation for asurface PLT of 150 mm × 150 mm (UOBCE)

1400

Estimated δ for 10 kPa

800
600

Mesured δ for 10 kPa

400
200

Measured δ for 6 kPa

Applied stress, q (kN/m2)

1200

0

Estimated δ for 6 kPa

0Measured δ for 2 kPa

1000

Estimated δ for 2 kPa

400

Estimated δ for 0 kPa

0
20

40

50

The correlation factor, f 1 value for estimating reliable
settlement behaviour of shallow footings in sands undersaturated condition is typically in the range of 4.5 to 5.0. On
the other hand, the correlation factor, f 2 value for estimating
the settlement of unsaturated sands falls between 4.5 and 7.5
for the sands evaluated. The need for using such a wide rangeof f 2 values (i.e., 4.5 to 7.5) can be attributed to the influenceof matric suction on the cone resistance, q c values which
contributes to a reduction of the settlement, of sands
under unsaturated conditions (i.e., (u a - u w ) > 0 kPa).

200

10

30

9 DISCUSSION OF RESULTS

Measured δ for 0 kPa

020

Fig. 12. Comparison between the measured and estimatedsettlements using the modified Schmertmann’s equation for a
FLT of 1500 mm × 1500 mm (FLT from [35])

800
600

10

Measured or estimated settlements, δ (mm)3040Measured or estimated settlement, δ (mm)Fig. 11. Comparison between the measured and estimatedsettlements using the modified Schmertmann’s [8] equationfor embedded PLTs of 150 mm × 150 mm (modified
UOBCE)It should be noted that the factors, f 1 and f 2 values for
saturated and unsaturated sand conditions respectively are
dependent on the relative density, D r . In both cases, thecorrelation factor increases proportionally with an increase in
the relative density, D r . These observations are consistent
with the conclusions drawn by researchers [34]. Estimatedand measured settlement values of large-scale footings (fourFLTs) from a reported in-situ case study presented in [35] arecompared in this study as shown in Fig. 13.
Because of the limited depth of the test box of both theUOBCE and the modified UOBCE in the laboratory, themaximum average matric suction value (i.e., (u a - u w ) AVR )simulated in the tank was 6 kPa. The proposed technique wasdeveloped based on test results using average matric suction

8.1 Validation of the Proposed Technique
Comparisons are provided between the estimated settlement(from the proposed technique) and measured settlement (fromin-situ FLTs conducted by [35]) values for both saturated andunsaturated sands to validate the proposed technique. Moredetails of the analysis and comparisons between the measured
and estimated settlement values using the proposed
relationships in Eq. (5) and measured settlement values forboth saturated and unsaturated sands using four in-situ
footing load tests (FLTs) are provided. The resultssummarized in Fig. 12 show the settlement values estimatedusing the proposed technique for a typical full-scale footing in
19

values ≤ 6 kPa; however, it provides good comparisons forfootings constructed in sands for higher matric suction values(i.e., 10 kPa) for the FLTs. The average matric suction value(u a - u w ) AVR for the FLTs in sand using [35] results wasdetermined assuming a constant matric suction distribution as
the depth of groundwater table is at 4.9 m and the natural
water content over this depth was constant.

Estimated settlement, δ (mm)30

The bearing capacity of unsaturated sands increases withmatric suction in a linear fashion up to the air-entry value(saturation zone). There is a non-linear increase in the
bearing capacity in the transition zone (i.e., from air-entry to
the residual suction value). The bearing capacity however
decreases in residual zone of unsaturation. Fig. 4 shows a setof typical results and behaviour of variation of bearing
capacity with respect to matric suction. The behaviour ofbearing capacity of unsaturated soils is consistent with theshear strength behaviour of unsaturated sands ([32], [33]).
Schmertmann’s equation [8] (i.e., Eq. (5)) with proposed
relationships for modulus of elasticity, E s for saturated andunsaturated conditions (i.e., Eq. (6) and Eq. (7)) provide
reasonably good comparisons between measured andestimated settlements for both model PLTs and in-situ FLTs
(see Figs. 10, 11, 12 and 13).
The modified Terzaghi and modified Schmertmann’s
procedures for predicting the bearing capacity and settlement
behavior of sandy soils, respectively for both saturated andunsaturated conditions are promising and can be used bypracticing engineers.

Measured = EstimatedMeasured vs estimatedsettlement for 4-full scale footings

25
20
15
10

B (mm) L (mm)1000
FLT 1000
FLT 1500
1500
FLT 2500
2500
FLT 30003000

511 ACKNOWLEDGMENTS
Funding received from Natural Sciences and Engineering
Research Council (NSERC) of Canada for this project is
acknowledged. The second author thanks the Ministry ofHigher Education and Scientific Research in Libya(M.H.E.S.R) for the financial support. The authors thank JeanClaude Celestin, technical officer of the GeotechnicalLaboratory, University of Ottawa for his assistance.

0
0

5

10

15

20

25

30

Measured settlement, δ (mm)Fig. 13. Comparison between estimated and measured
settlements of four full-scale footings (FLTs from [35]) in
unsaturated sands corresponding to different applied stresses
10

12 REFERENCES[1] Terzaghi K, “Theoretical Soil Mechanics,” John Wileyand Sons. New York, NY, USA, 1943.
[2] Meyerhof GG,“The ultimate bearing capacity offoundations,” Géotechnique, 1951, (2), 301 – 332.
[3] Vanapalli SK and Mohamed FMO, “Bearing capacity ofmodel footings in unsaturated soils,” In Experimental
Unsaturated Soil Mechanics, Springer Proc. in Physics.
Springer-Verlag Berlin Heidelberg, 2007, 112, 483 –493.
[4] Vanapalli SK, “Shear strength of unsaturated soils andits applications in geotechnical engineering practice,”
Keynote Address. Proc. 4th Asia-Pacific Conf. on
Unsaturated Soils. New Castle, Australia, Edited by O.
Buzzi, S. Fityus and D. Sheng. 2010 Taylor & Francis
Group, London, ISBN 978-0-415-80480-6, 2009, 579 –598.
[5] Oh WT and Vanapalli SK, “Modeling the stress versus
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[6] Oh WT and Vanapalli SK, “Modelling the settlement
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[7] Mohamed FMO, Vanapalli SK and Saatcioglu M,“Settlement estimation of shallow footings in saturated
and unsaturated sands using cone penetration test (CPT)SUMMARY AND CONCLUSIONSThe bearing capacity and settlement behaviour of sandy soil
under saturated and unsaturated conditions using surface andembedded model footings tests are studied in this research
program. The bearing capacity values are underestimated forsurface model footings (i.e., the depth of the model footing isequal to zero) when calculations are based on effective
friction angle, ' = 35.3 º for the tested sand both insaturated and unsaturated conditions. Typical value of
dilatancy angle for sands is equal to 10 % of effective friction
angle, ' (see Table 3). Reasonably good comparisons wereobserved between measured bearing capacity and predictedvalues using Eq. (2) particularly when the influence of thedilatancy angle,
 account.
towas taken in
There is no need to increase the effective friction angle, 'by 10% to obtain reasonable comparison between themeasured and estimated bearing capacity values (see Table 4)
for embedded footings. In other words, the dilatancy angle,
 = 0 for shallow foundations whose D f /B ≈ 1. Such acontrasting behavior between surface and embedded footings
may be attributed to the contribution of the overburden stresswhich eliminates the influence of dilation. These observations
are consistent with the conclusions drawn in ([19], [23] and[31]) with respect to dilation effects in sandy soils.

20results,” 2011 CSCE Annual Conf. – Engineers
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[8] Schmertmann JH, Hartman JP and Brown P, “Improved
strain influence factor diagrams,” J. of GeotechnicalEng. Division, ASCE, 1978, 104 (8), 1131 – 1135.
[9] DeBeer E, “Bearing capacity and settlement of shallow
foundations on sand,” Proc. of Symposium on Bearing
Capacity and Settlement of Foundations, Durham, NC,
1965, 15 – 33.
[10] Broms BB, “The effect of degree of saturation on the
bearing capacity of flexible pavements,” Highway
Research Record, 1963, 71, 1 – 14.
[11] Steensen-Bach JO, Foged N and Steenfelt JS, “Capillary
induced stresses – fact or fiction?,” 9th ECSMFE,
Groundwater Effects in Geotechnical Engineering,
Dublin, 1987, 83 – 89.
[12] Fredlund DG and Rahardjo H, “Soil mechanics forunsaturated soils,” 1st ed., Wiley, New York, 1993.
[13] Costa YD, Cintra JC and Zornberg JG, “Influence of
matric suction on the results of plate load tests
performed on a lateritic soil deposit,” Geotechnical
Testing J., 2003, 26(2), 219 – 226.
[14] Oloo SY, Fredlund DG and Gan JK M, “Bearing
capacity of unpaved roads,” Canadian Geotechnical J.,
1997, 34, 398 – 407.
[15] Vanapalli SK and Oh WT, “Mechanics of unsaturated
soils for the design of foundation structures,” Plenary
Address. Proc. 3rd WSEAS International Conference on
Engineering Mechanics, Structures, Engineering
Geology, Corfu Island, Greece, 2010, 363 – 377.
[16] Mohamed FMO and Vanapalli SK,“Laboratory
investigations for the measurement of the bearing
capacity of an unsaturated coarse-grained soil,”
Proceedings of the 59th Canadian Geotechnical Conf.,
Vancouver, B.C, 2006, 219 – 226.
[17] Meyerhof GG, “Penetration tests and bearing capacityof cohesionless soils,” J. of Soil Mechanics andFoundation Division, ASCE, 1956, 82 (1), 1 – 19.
[18] Das B and Sivakugan N, “Settlement of shallow
foundations on granular soil – an Overview,” Int. J. of
Geotechnical Eng., 2007, 1, 19 – 29.
[19] Bolton MD, “The strength and dilatancy of sands,”
Géotechnique, 1986, 36 (1), 65 – 78.
[20] Sfriso A, “The friction angle and critical state void ratioof sands,” Proc. of 17th Int. Conf. of Soil Mechanics andGeotechnical Eng., ICSMGE, Alexandria, Egypt, 2009,433 – 435.
[21] Vermeer PA and DeBorst R, “Non-associated plastic forsoils, concrete and rock,” Heron, 1984, 5 – 64.
[22] Bishop AW, “Shear strength parameters for undisturbed
and remoulded soils specimens,” In stress strain
behaviour of soils. Foulis, London, 1972, 3 – 58.
[23] Chakraborty T and Salgado R, “Dilatancy and shear
strength of sand at low confining pressure,” J. of
Geotechnical and Geo-environmental Eng., 2010, 136,
527 – 32.
[24] Poulos H and Davis E, “Elastic solutions for soil androck mechanics,” John Wiley and Sons, NY, USA,1974.

[25] Mohamed FMO and Vanapalli SK, “An experimental
investigation of the bearing capacity of an unsaturated
sand using cone penetration tests,” Proc. 62nd Canadian
Geotechnical Conf., Halifax, NS, 2009, 205 – 213.
[26] Kumbhokjar AS, “Numerical evaluation of Terzaghi’sN ,” J. of Geotechnical Eng., American Society of Civil
Engineers, 1993, (3), 598 – 607.
[27] Hansen BJ, “A revised and extended formula for bearing
capacity,” Bulletin No. 28, Danish GeotechnicalInstitute, Copenhagen, Denmark, 1970, 5 – 11.
[28] DS 415, “The Danish code of practice for foundationengineering,” Danish Society of Civil Engineering,
1984.
[29] Mohamed FMO, “A semi-empirical approach for theinterpretation of the bearing capacity of unsaturated
soils,” Master’s Thesis, University of Ottawa, Ottawa,
ON, Canada, 2006.
[30] Liao HJ and Hsu ST, “Uplift
behaviorofblade-underreamed anchors in silty sand,” J. of
Geotechnical and Geoenvironmental Eng., 2003, 560 –
568.
[31] Vesić AS and Clough GW, “Behaviour of granularmaterials under high stresses,” J. of Soil Mechanics andFoundation Divison. ASCE 94, 1968, SM3, 661 – 688.
[32] Vanapalli SK, Fredlund DG, Pufahl DE and Clifton AW,“Model for the prediction of shear strength with respect
to soil suction,” Canadian Geotechnical J., 1996, 33 (3),379 – 392.
[33] Vanapalli SK and Lacasse F, “Comparison between themeasured and predicted shear strength behavior of four
unsaturated sands,” Proc. of the 4th Asia-Pacific Conf.
on Unsaturated Soils. New Castle, Australia. Edited by
O. Buzzi, S. Fityus and D. Sheng. 2010 Taylor &Francis Group, London, ISBN 978-0-415-80480-6,
2009, 121 – 127.
[34] Lee J and Salgado R, “Estimation of footing settlementin sand,” The Int. J. of Geomechanics, 2002, 1 (2), 1 –28.
[35] Giddens R and Briaud J, “Load tests on five large spreadfootings on sand and evaluation of prediction methods,”
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21

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Development of Piled Geo-wall(New type reinforced soil wall)
Hara, T., Tsuji, S., Yoshida, M. and Sawada, K.
Gifu University, Japan

ABSTRACTThis study aims to achieve a new type of independent reinforced soil wall (Piled Geo-wall) that can be substitute forone made up of concrete with similar scale, and to contribute to sustainable development. In order to confirm thepracticability of the novel structure, three experimental studies with static and impact loading tests and a dynamic
centrifuge model test were carried out in the past years, which were reported on the 1st GEOMATE conference.
Simple design methods of the novel structure, which can reproduce the results of the experiments, are introduced inthis paper.
Keywords: Reinforced soil wall, Pile foundation, Earth-retaining wall, Rock-fall protection wall
Corroboration of Pile andreinforced soil wall

1. INTRODUCTION
The high ductility of soil structure reinforced by geogrid iswell known, as is the possibility of building independent soil
wall. The independent reinforced soil wall has been applied
to such diverse structures as rock-fall protection walls, mudand snow avalanche protection walls and the suchlike. Since
it can be built using existing soil at the construction site if it
is compactable one, they are being used ever more
frequently as one of economic and CO2 reducible structures.
At present, however, the adoption of the spread foundation
for the independent reinforced soil wall makes the design too
wide for application to narrow construction sites, such as
beside mountainous road. If a narrow independent reinforced
soil wall as like as RC wall with pile foundation is achieved,
it could be widely applied. And it can also be substitute forconcrete ones and contribute sustainable development.
Therefore, a new type of independent reinforced soil wallwith inserting piles into the Geo-wall body, as shown in Fig.
1, which is referred to “Piled Geo-wall or PGW” in thispaper, has been developed. The practicability of Piled
Geo-wall to diverse structures has been already confirmed
from three experimental studies with full-scale static andimpact loading tests and a dynamic centrifuge model test
(25G) were carried out in the past years [1], [2].
Recognizing the practicability, a study on simple design
method to apply Piled Geo-wall to earth-retaining walls,
rock-fall protection walls, and seismic countermeasure wallsof road embankment in order to reduce deformation of roadsurface, has been conducted. In this paper, because of paper
limitation, simple external stability verification methods ofPiled Geo-wall as well as the reproducibility of theexperiments’ results by using the proposed verification
methods are introduced.

(Piled Geo-wall)Fig. 1. Image of Piled Geo-wall (PGW).
2. FUNDAMENTAL EXTERNAL STABILITY MODEL
Piled Geo-wall was devised from an assumption regarding
the interaction between pile and Geo-wall body, which the
Geo-wall body with high ductility is possible to transmit
lateral forces to the piles despite large relative displacement
between the pile and the Geo-wall body as shown in Fig.2.
Hence, in the design, the responses of the pile and theGeo-wall body have to be estimated respectively. Fig. 3
shows the proposed simple estimation model of the
responses of the pile and the Geo-wall body of Piled
Geo-wall.
Force
P

Pile foundation

Soil structure(with geogrid)
Soil structure(without geogrid)
Displacement

Fig. 2. Interaction image between pile and Geo-wall body.

22

B
b

b

Earth pressure
etc.

EGAG
EG I GEGAG

EP I P
EPAP

KSG
KVG

KRG
kH

KVP

(a) Estimation model for pile response

perfectly plastic vertical spring constant beneath the pile
(kN/m) set from Eq. (6), the limited value of KVP is theultimate bearing capacity of pile, kh and kv: coefficients ofhorizontal and vertical subgrade reaction (kN/m3) in normal
time or seismic situation, which are estimated fromSpecifications for Highway Bridges (Part IV) [3], D: pile
diameter (m), du: unit depth, namely 1.0m, AG and AP: crosssection areas of the Geo-wall body and the pile (m2), c andφ: cohesion and shear resistance angle of foundation ground(kPa and rad).
K H = kh ⋅ D ⋅ d u(1)

KVG = kv ⋅ AG

(2)

K SG = 1 3⋅ kv ⋅ AG

(3)

Smax = c ⋅ AG + σ ⋅ tan φ(4)

K RG = kv ⋅ b(5)

KVP = kv ⋅ AP

(6)

B
b

b

Earth pressure
etc.

EGAG

3. PRACTICABILITY OF THE MODEL TO DESIGN
3.1 Earth retaining wall in normal time design situation

EG I GEGAG
EG I P
EGAP

Rigid body
KSG

The summary of the static loading test of Piled Geo-wall [1],which is targeted for verification of the practicability of the
proposed external stability model to the design of earth
retaining wall in normal time situation, is shown in Fig. 4.
Cross section1000 500

KVG

EP I P
EPAP

kH



 

2000

KRG

H200X200
Jack (50t)
Load cell

Reaction wall
Plate
EPS

Where, EPIP, EGIG: flexural stiffness of pile and Geo-wall(kNm2/m), EPAP, EGAG: compressional stiffness of pile andGeo-wall (kNm2/m), EGIP: extent without considering
flexural stiffness of pile, kH: elasto-perfectly plastic spring
constant of horizontal subgrade reaction (kN/m) set from Eq.
(1), the limited value of kH is passive earth pressure, KVG:
elasto-perfectly plastic spring constant of vertical subgrade
reaction beneath the Geo-wall body (kN/m) set from Eq. (2),the limited value of KVG is ultimate bearing capacity, KSG:
elasto-perfectly plastic spring constant of horizontal shear
reaction beneath the Geo-wall body (kN/m) set from Eq. (3),the limited value of KSG, Smax, is set from Eq. (4), KRG:
elasto- plastic rotation spring constant beneath the Geo-wallbody (kN/rad) set from Eq. (5), the limited value of KRG isdetermined by the ultimate bearing capacity, KVP: elasto-

Longitudinal
additional
geogrid
H300X300Plane view  EPS

Plate
H200X200
Reaction wall

3000(b) Estimation model for Geo-wall body response
Fig. 3. Fundamental external stability model.

4000

KVP

Load cellFig. 4. Summary of target static loading test.

 

Fig.5 shows the results of dynamic penetration tests
converted to N value of standard penetration test (SPT) as
the ground condition of the test. In this verification, theresults of PD-2 obtained at the vicinity point of the target
Piled Geo-wall was referred.
N-valueof SPT (converted)
１
0cm 毎の換算Ｎ
d値 （回）
0

10

20

30

40

3.2 Earth retaining wall in seismic design situation
Fig. 8 shows the transverse section of the model adopted in
the dynamic centrifuge test (25G) [2], which is targeted forverification of the practicability of the proposed externalstability model to the design of earth retaining wall inseismic situation. In this test, Piled Geo-wall was adopted asa seismic countermeasure for prevention of large
deformation of the road embankment built on a slope. Slopground was made up of cement-stabilized soil and, the soilof the embankment and the Geo-wall body was compactedwith density control. Geotechnical and structural parametersconverted to actual scale are presented on Table.1 and 2,
respectively. Fig. 9 shows the input earthquake wave
converted to actual scale.

50

0
PD -1

深
さ （G L-	
 m ）
Depth

1

And Fig. 7 shows comparisons with the analysis results ofmaximum response of pile and measured ones from the test.
According to the results, it is confirmed that the proposedmodel can reproduce the response of Piled Geo-wall andrelative displacement between pile and Geo-wall body.

PD -2

2
3
4
5
6

Fig. 5. Ground condition.
Fig. 6 shows the analysis results of the relationship between
static load and displacement of Piled Geo-wall, top of thepile and Geo-wall body, which are compared with measured
value from the test.

Fig. 8. Model of dynamic centrifuge test (25G).







800



600

 
 




2
Acceleration(cm/s
Acceleration(cm/s))

 















6

5

5

4

4



Pile length (m)Pile length (m)

 
2

 

0
0

10

20

30

40

50

Displacement (mm)

30

-800

Fig. 9. Input earthquake.
Table. 1. Geotechnical parameter.
E0 (kPa)c (kPa)Geo-wallφ (deg)3.26 x 105

550

3.0 x 104040

3.0 x 104040

E0: deformation modulus, c: cohesion, φ: shear resistance angle



0

20
time (sec)-600

Embankment

2
1

10

-400

Slope

3

1

0

-1000



3

200-200 0Fig. 6. Relative displacement between pile and Geo-wall.
6

400

Table. 2. Structural parameter.
E (kPa)
0

50

100

150

Bending Moment (kNm)

Fig. 7. Maximum response of pile.

A (m2/m)

Pile

2.0 x 108

-34.79 x 10

Geogrid

8.0 x 105

1.0 x 10-3

I (m4)
2.04 x 10-4

E: Young modulus, A: sectional area, I: moment of inertia-

Because recognizing of dynamic interaction between
structure and surrounding ground is necessary in order todesign the structure by static analysis, dynamic interaction
between the Piled Geo-wall and the embankment, transitionof pile response and earth pressure at the embankment sideof the Piled Geo-wall at the time that the maximum pile
response was obtained are shown in Fig. 10.

0.58 m/sec2

4.7 m/sec2

	
 
	
  	
 

0$
!3$
!6$

Embankment$!9$

4.7 W2 / g

W2

6

6

5

5

4

4

 


6.5$

7$

Time (s)

7.5$

8$

(a) Dynamic interaction between PGW and embankment
150%

Pile length (m)Geo!wall$body$(10th$layer)$!15$

40 kPa

Geo!wall$body$(5th$layer)$!12$

Strain'(µm)

W1(a) Dynamic interaction(b) Load situation
Fig. 11. Design situation.

6$

Pile length (m)Accelera'on*(m/s2)

9$
3$

0.58 W1 / g

4.8 m/sec2

3
 
2

3
2



0%

1

1

!150%

	
 

!300%

0

0
0

!450%
6.5%

7%

Time (s)

7.5%

10

20

30

40

50

Displacement (mm)

8%

0

50

100

150

Bending Moment (kNm)

Fig. 12. Maximum response of pile.

(b) Transition of maximum pile strain
Earth&pressure&(kPa)

120"
100"
80"

	
 

60"
40"
20"
0"
6.5"

7"

7.5"

8"

Time (s)

(c) Transition of earth pressure
Fig. 10. Dynamic responses.
According to the results, the antiphase between the responseof Geo-wall body at 5th layer (the part with inserting pile)and ones of embankment and Geo-wall body at 10th layer(the part without inserting pile) is confirmed. And the statesof inertia forces and earth pressure acting on theembankment side of Piled Geo-wall body at the time that the
maximum pile response was obtained were confirmed asshown in Fig. 11. Fig.12 shows comparisons of the
maximum response of the pile, which analyzed one by theconfirmed load states and obtained one from the experiment.
From this result, it could be confirmed that the proposed
estimation model can reproduce well the actual response of
the pile of the Piled Geo-wall.

In practical design, however, considering real dynamic
interaction and load situation on the design stage is difficult.
The inertia forces are estimated from design horizontal
seismic coefficient of 0.8 in the experimental groundcondition and the earthquake level and the earth pressure
with the coefficient of earth pressure predicted by Eq. (7) is
considered in Japan [4]. Large load situation is assumed in
the design as the result as shown in Fig. 13.
Fig. 14 shows the maximum pile response analyzed byproposed model with the deign load situation shown in Fig.
13. An abnormal bending moment is occurred at the upper
part of the pile as well as larger response than real one is
obtained. This is caused by the considering of excessive
inertia force, about 14 times of real one, at the upper part of
the Geo-wall body (the part without inserting pile). From thefollowing viewpoints, however, it is considered that the
proposed model can be applied to practical design.
- In the design executed on the stage that the real dynamic
interaction and the real load situation, overestimating the
response of the target structure is practical rather than the
reproduction of actual one. It is the practical design.
- From specifying that the piles have to be inserted to the topof Geo-wall body, occurrence of abnormal bending
moment at upper part of the pile can be prevented.

K EA = 0.26 + 0.97 ⋅ kh

(7)

Where, KEA: coefficient of active earth pressure during

earthquake, kh: design horizontal seismic coefficient
0.8 W1estimated by the proposed model and ones obtained from the
dynamic centrifuge model test (25G) shown in Fig.8. The
measured points of the road surface deformation in theexperimentare shown in Fig. 16.
W1
60 kPa0.8 W2W2

Measuredpoints

PGW

Fig. 13. Design load situation.
6

6

5

5




























Road surface

Fig. 16 Measured points of the road surface deformation.

4

Pile length (m)Pile length (m)

3
 
2

4
3
2


1

1

0
-100

0
0

10

20

30

40

50

0

100

200

300

Bending Moment (kNm)

Displacement (mm)Fig. 14. Maximum response of pile.
Fig. 15 shows the comparison with the analysis result of the
maximum response of Geo-wall body and measured one
from the experiment. According to the results, it isconfirmed that the proposed model can reproduce the real
response of the Geo-wall body.

Road surface deformation
Road surface
Input estimated responseof Geo-wall body4

Geo-wall height (m)The road surface deformation depending on maximum
response of Piled Geo-wall estimated by the proposed modelis analyzed by using simple two-dimensional static elastic
FEM as shown in Fig. 17. In this analysis, road surface
deformation is analyzed by inputting estimated maximum
Geo-wall response from the proposed model as an imposed
deformation.
Fig. 18 and 19 show the entire deformation results obtained
from FEM and the comparison with analyzed road surface
deformation by FEM and measured one from the experiment.
According to the results, the analyzed deformation
reproduces measured one comparatively well, hence it canbe confirmed that the proposed model can apply to theverification of seismic performance of road embankment
with Piled Geo-wall as a seismic countermeasure.

3

	
 
	
 

2

Fig. 17. Analysis of road surface deformation.

1

 
Subsidence
0
0

50

100

150

200

Displacement (mm)Fig. 15 Maximum response of Geo-wall body.
3.3 Seismic countermeasure wall for road embankment
The practicability of proposed model to the design on
seismic countermeasure wall for prevention of large
deformation of road embankment is confirmed fromcomparison with analyzed road surface deformation
depending on maximum response of Piled Geo-wallFig. 18. Entire deformation obtained from FEM.

Fig. 22 shows proposed conversion procedure of design
static load from impact-load in this study.

Subsidence (mm)

0
100mm

-20

   
  

-40
-60
-80
-100
-2

0

START

Scale ofhorizontal
displacement246Distance from PGW (m)Estimation of maximum
static load (Pmax) by halfsine wave method

Energy absorption
by Geo-wall body

8

Estimation of load
reduction by plastic
deformation of Geo-wall

10

Energy absorption
by pileEstimation of natural
period of PGWFig. 19. Road surface deformation.

Estimation of load
response ratio to pile

3.4 Rock-fall protection wall
The impact-loading test of Piled Geo-walls [1], which istargeted for verification of the practicability of the proposed
external stability model to the design of rock-fall protection
wall, is shown in Fig. 20. In this verification, the results of
two impact-loading tests to actual scale models, one isadopted in the static loading test (PGW-1) shown in Fig.4
and another is a new one (PGW-2), are targeted. The piles of
PGW-2 are installed at outside of Geo-wall as shown in Fig.
21, are adopted.

Estimation of load
reduction by un-linear
behavior of pileor 
Iron ball (9kN)
0.9tf
50100kJ

Geo-wall

Where, the maximum static load (Pmax) based on half sine
wave method is estimated by Eq. 8. In this equation, m is
mass, V0 is initial velocity (m/sec) at the time of therock-fall impact to the Geo-wall body, which is predicted by
Eq. 9, T is sustaining period (sec) of the impact by PGW,
which is assumed as 0.07sec from the experiments, g is
G-force, h is falling height of the rock fall.

Fig. 20. Impact-loading test.

π ⋅ m ⋅V0
T

(8)
(9)

V0 = 2 ⋅ g ⋅ h

3.0m

Elevation

END

Fig. 22. Conversion procedure of design static load.

Pmax =

11.1m

2.0m

100kJ impact

Determination of“Design static load” Energy

Cross section

absorption by Geo-wall body was assumed asfollows;
a. The displacement-time relationship of the iron ball andthe pile obtained from the experiment is converted to theload reduction-time relationship as shown in Fig. 23. The
maximum load reduction is the maximum static load
estimated by Eq. 8. Load reduction by pile is assumed
from inverse analysis with using proposed externalstability model, because high reproducibility of the static
loaded pile displacement by using the proposed model
was confirmed at the section “3.1”.

H300X300

600

Overall

Load (kN)

500

Plane view
：strain gauge (geogrid)：strain gauge (pile)

Pile

400

Absorbed energy by Geo-wall

300
200
100

Absorbed energy by pile

0
0

0.01

0.02

0.03

0.04

0.05

0.06

Time (sec)Fig. 21. Structural condition and measurements of PGW-2.

Fig. 23. Load reduction-time relationship0.07

b. The iron ball displacement expresses entire energy
absorption, thus the difference of entire load reduction
and one of pile was assumed as the reduction load byplastic deformation of Geo-wall body.
c. From the abovementioned result, the relationship between
load reduction ratio and plastic displacement of Geo-wallbody was assumed as Eq. 10.

PR (kN)(10)

δ = Pmax K s(11)

PR max = Pmax ⋅ (1− cRP )

(12)

100

50

0
0

10

20

30

Displasment (mm)Fig. 24. An example of equivalent energy method.

cres = −4 ⋅ rt ⋅ ( 0.6 ⋅ rt −1.1)(13)

rt = T t(14)

PRES = PR max ⋅ cres

(15)

Where, cres: load response ratio to pile, rt: the ratio of impact
period of road and natural period of pile, as Eq.14, t: natural
period of pile (sec), PRES: load transmitting to pile (kN).
And finally, the design static load (PD) is determined fromthe load reduction, which is estimated by equivalent energy
method [6] of both elastic and elasto-plastic analyses withusing the proposed external stability model. Fig. 24 shows
an example of equivalent energy method applied to PGWdesign.

Fig.25 and 26 show the maximum pile responses of PGW-1
and 2, respectively. According to the results, it can be
confirmed that the proposed external stability model
reproduces well ones obtained from the experiments. Where,
although it is no wonder that analyzed pile top displacements
match up to obtained ones because the energy absorption of
Geo-wall body was determined from inverse analysis of pile,the reproducibility can be confirmed from the distribution in
depth of the displacement and the bending moment of piles.
6

6

5

5

4

4

 

Pile length (m)

3
 
2

3
2


1

1

0

0
0

10

20

30

0

40

50

100

150

Bending Moment (kNm)

Displacement (mm)Fig. 25. Maximum pile response (PGW-1).
6

6

5

5

4

4

 


Pile length (m)Where, cRP: load reduction ratio caused by plastic
deformation of Geo-wall body, δ: plastic deformation of
Geo-wall body (m) estimated by Eq. 11, Ks: equivalent
plastic spring constant with respect to inserting rock-fall into
Geo-wall body (kN/m), which is assumed as 1650kN/m
from the experiments, PRmax: reduced static load withconsidering energy absorption by Geo-wall.
The values of the impact period (T) and the equivalent
plastic spring constant (Ks) are considered as variables
depending on stiffness of the Geo-wall body and so on, but
because the stiffness would not be so large different one if it
is similar scale PGW with experiment one, it is consideredthat the values can be applied to small type of PGW.
Load response ratio to pile expresses reduction (oramplification) effect of transmitting load to pile. It is wellknown that the response reduces if the structure with longnatural period receives the impact load with short impact
period; in contrast, the response amplifies if the structurewith short natural period receives the impact load withcomparative long impact period. Therefore, the coefficientconsidered the characteristics in case of using load estimatedfrom half sine curve method was proposed, as Eq.13, fromthe past study [5]. According to the past study, the border of
reduction and amplification of the response is said to be thecase that the ratio of impact period of road and natural
period of structure is about 0.26.

Pile length (m)(0.15 ≤ δ )Liner Response
Design PD = 128 kN

Pile length (m)cRP = 9.8⋅ δ ⋅ (δ − 0.144)

PRES = 150 kN

Elasto-Plastic
150

3
 

2

3
2


1

1

0

0
0

10

20

30

40

Displacement (mm)

0

50

100

150

Bending Moment (kNm)

Fig. 26. Maximum pile response (PGW-2).

4. CONCLUSIONSThe contents of this paper are concluded as follows;- Simple external stability model of Piled Geo-wall forpractical design was proposed.
- The practicability of the proposed design model was
confirmed from good reproducibility of real response
obtained from full-scale static and impact loading tests,
and a dynamic centrifuge model test.
- In the design on Piled Geo-wall in static load situation,high reproducibility of interaction between pile andGeo-wall body as well as one of pile response wereconfirmed.
- In the design on Piled Geo-wall in seismic situation, it was
confirmed that safety design can be executed in case withdesign load situation as well as high reproducibility of thepile response in case with real load situation.
- In the design on Piled Geo-wall for rock-fall protection,
conversion procedure of design static load from impact
one was also proposed and its practicability was confirmed
from good reproducibility of pile response by using the
proposed procedure.
The following issues, however, have to be conducted toapply the proposed model to more wide condition of Piled
Geo-wall.
- Application of the method to design on large scale (width,height) Piled Geo-wall.
- Application of steel-pipe pile and multiple rows’ piles toPiled Geo-wall.
- Improvement of static load conversion procedure from
impact one as being able to apply to diverse conditions of
Piled Geo-wall.
Meanwhile, simple internal stability model has been alsostudied; it will be presented after more verification ofvalidity.
REFERENCES[1] Hara T, Tsuji S, Yoshida M, Ito S and Sawada K,“Experimental development of new type reinforced soil
wall,” Int. J. of GEOMATE, June 2012, Vol. 2, No.2 (Sl.
No.4), pp.213-218[2] Hara T, Tsuji S, Yashima A and Sawada K,“Independent reinforced soil structure with pilefoundation,” J. of SOILS AND FOUNDATIOS, Vol.50,
No.5, 2010, pp. 565-571.
[3] Japan Road Association (JRA), “Specifications ofhighway bridges (Part IV: Substructures), Mar. 2002
[4] Japan Road Association (JRA), “Specifications ofhighway bridges (Part V: Seismic design), Mar. 2002
[5] Clough RW and Penzien J, “Dynamics of structures,”
McGraw-Hill, 1975, p.80.
[6] Newmark NM and Veletson AS, “Effect of inelastic
behavior on the response of simple systems toearthquake motions,” Proc. of 2nd WCEE, 1960,
pp.895-912

29

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Measurement of Strain Distribution Along Precast Driven Pile During
Full Scale Pile Load Test.
Faisal Ali and Lee Sieng Kai
Department of Civil Engineering, National Defense University of Malaysia

ABSTRACTOne of the basic parameters required for forecasting pile deformation under loads is the characteristics and profile of the
shaft load transfer from the pile to the surrounding soil. This parameter can be obtained my measuring the strain
distribution along the pile during full-scale pile load test. For cast-in situ bored pile the strain measurement can be easily
done by fixing the sensors to the reinforcement cage before pouring the concrete. For precast driven / jack-in piles, theapplication of instrumented full-scale static load tests is far more challenging than their bored pile counterparts due tosignificant difference in method of pile installation. Due to practical shortcoming of conventional instrumentation
method instrumented full-scale static load tests are in fact rarely used in driven pile application in this region. For hollow
precast spun concrete piles attempts have been made by geotechnical engineers to measure the strain distribution by
installing either an instrumented reinforcement cage or instrumented pipe into the hollow core of spun piles followed bycement grout. However in this method, known as approximate method, the measured characteristics cannot be
considered to represent the working piles because the presence of reinforcement/pipe and the grout would alter the
stiffness of the test pile. In this paper a method to measure the strain distribution along the installed precast spun concrete
pile during full-scale pile load test is described. The method utilizes the hollow core of the spun pile without changing thephysical properties e.g. stiffness, of the pile. The main advantages of this method are: the measurement can be done at
any location along the pile, the sensors can be retrieved and the measured characteristics are representative of those of theworking piles.
Keywords: pile, ultimate capacity, instrumentation, spring-loaded transducer
base; (d) Pile dimensions; and (e) Stiffness of the pilematerial.
This recent development in the understanding of the loadtransfer and bearing behaviour of piles in fact exerted asignificant and positive influence on the evolution of codes ofpractice and design methods for foundations in somecountries. For example, the revised Singapore Standard on
Code of Practice for Foundations CP4: 2003[3], recommends
that the static load test on preliminary test pile be
instrumented to measure the transfer of load from the pile
shaft and pile toe to the soil. The Code also recommends thatfor driven piles (similar to bored cast-in-place piles), the axial
load capacity can be evaluated empirically from correlation
with standard penetration tests SPT N-values (which arewidely used in this region) using modified Meyerhof
Equation, where the ultimate bearing capacity of a pile incompression is given by:

1. INTRODUCTION
The high strength prestressed spun concrete piles, commonly
driven with hydraulic impact hammers or preferably installedwith jacked-in rigs when considering the stricter regulations
with respect to noise and vibrations in more urban areas, often
offer a competitive choice of foundation system for projects
with medium and high loadings. They are widely used infoundations for schools, high-rise buildings, factories, ports,
bridges and power plants in this region.
In early years, the main construction control for driven piles
was mostly based on the measurement of set of each pile
coupled with a selected small number of non-instrumented
static load tests to verify the specified load-settlement
requirements.
In recent years, with critical evolution in the understanding of
the load transfer and bearing behaviour of piles mainly
through analysis of instrumented full-scale load tests(particularly for bored cast-in-place piles), many engineers
can now appreciate that the pile performance is not simply amatter of ultimate load value alone. According to Fleming[1][2] some of the basic parameters required for forecasting
pile deformation under loads include (a) Ultimate shaft load
and its characteristics of transformation to the ground; (b)Ultimate base load; (c) Stiffness of the soil below the pile

Qu = Ks*Ns*As + Kb*(40Nb)*Ab(1)

where:
Qu = ultimate bearing capacity of the pile, kN,
Ks = empirical design factor relating ultimate shaft load to
SPT values, kN/ m2 per SPT blow,
Ns = SPT value for the pile shaft, blows/300mm,
As = perimeter area of the shaft, m2,
30

Kb = empirical design factor relating ultimate end bearing
load to SPT values, kN/ m2 per SPT blow,
Nb = SPT value for the pile base, blows/300mm, and
Ab = cross-sectional area of the pile base, m2.

N (blows/30cm)
0

50

100

150

0

(Pile head)
Existing Ground Level
Strain Gauges Lev. A

5For bored piles, instrumentation using sacrificial cast-in
vibrating wire strain gauges and mechanical tell-tales which
permit for monitoring of axial loads and movements at
various levels down the pile shaft including the pile toe level
had been practiced successfully within limits of accuracy
posed by constraints inherent of the installation method, in
this region for many decades, allowing insight evaluation ofKs and Kb factors, including well documented history of“The Design of Foundations for Suntec City, Singapore”[4][5]. Recent use of Global Strain Extensometers (Glostrext)
Method for bored piles in this region [6][7] also provides an
excellent alternative means for similar purpose, but with thecapability of producing a more reliable performance andaccurate results. For precast driven piles, the application of
instrumented full-scale static load tests is far morechallenging than their bored pile counterparts due tosignificant difference in method of pile installation. Due topractical shortcoming of conventional instrumentation
method and the lack of innovation in this area, instrumented
full-scale static load tests are in fact rarely used in driven pileapplication in this region. Therefore, the far lacking driven
pile industry is long due for a better technology to
revolutionize the methodology in the acquisition of design
data in a more accurate and reliable way, to catch up with theevolution in the design methods.

Clay

Hollow annular space
Spun Pile

10

Strain Gauges Lev. B
Depth (m)

15

Sandy
Clay

Strain Gauges Lev. C

20

Strain Gauges Lev. D
25

Sandy
Silt

30

SI borehole log

Strain Gauges Lev. E
Strain Gauges Lev. F
Pile toe at 30.0 m depth
Legends:
denotes high temperature - resistant
Strain Gauges
denotes Pile Joint

Fig. 1 - Diagrammatic illustration of conventional spun pile
instrumentation scheme
2.2 Approximate Instrumentation Method
Due to the difficulties of using the conventional method, theengineering community for spun pile industry has been using
an approximate instrumentation method for the past few
decades, by installing either an instrumented reinforcement
cage or an instrumented pipe, into the hollow core of spunpiles followed by cement grout infilling (Fig. 2(a)).
As this approximate method is comparatively more“convenient” to be implemented than the conventional
method, it was widely practiced in this region for the past few
decades. Some contract specifications also ask for the
inclusion of conventional sleeved rod extensometers(depending on the space available) to monitor the pile
shortening reading during the static load tests. Either using an
instrumented reinforcement cage or an instrumented pipe,with or without the added-in sleeved rod extensometers, theend product after the cement grout infilling is more towards asolid pile, as shown in Fig. 2(b).
Therefore the obvious shortcomings of this approximate
method include:(a) The infilling of cement grout substantially alters the
structural properties of the piles, thus rendering themsignificantly different from the actual working spunpiles, which are usually not grouted internally;(b) The presence of the grout would change the overall
stiffness of the pile and therefore the measured load
distribution along the pile would be different under theapplied loading.

2. CURRENT METHODS OF PILE
INSTRUMENTATION
2.1 Conventional Instrumentation Method
A conventional instrumentation scheme for spun pile staticload testing is shown in Fig. 1. The method involves
incorporating high temperature-resistant strain gauges into
the heat-cured production process of prestressed spunconcrete piles.
This method is extremely unpopular and difficult to beroutinely applied in project sites due to the following
constraints:(a) High cost of these temperature-resistant strain gauges;(b) Tremendous difficulties involved in coordinating the
installation of the strain gauges into pile segments;(c) Long lead-time is normally required for instrumentation
works, as the instruments have to be pre-assembled andinstalled onto the high strength prestressing bar cageprior to heat-cured ‘spin-cast’ production process of thepiles; and(d) Great uncertainty over the ability of the delicate
instruments to withstand the stresses arising from pile
production and driving processes.
31

(c) Structural shortening measurement of the test piles are
not representative of the actual working piles;(d) Structural integrity of the original pile cannot be reliably
ascertained, particularly performance of pile joints,
during the static load test as the stiffness of the pile has
been altered; and(e) Significant time loss due to grout infilling and curing
process, beside the environmental unfriendly nature ofthis method.
N (blows/30cm)
0

50

100

150

0

(Pile head)

3. RECENTLY DEVELOPED INSTRUMENTATION
To address the challenges and difficulties posed by theconventional and approximate methods, Global Strain
Extensometer technology for spun piles had been developed,
improved and field tested.
In order to meet the requirements for practical
instrumentation application on spun piles, the desired
characteristics for the retrievable pneumatically anchoredextensometer system were identified as follow:(i) To utilize the hollow core of spun piles (or cylinder
concrete piles) as an ideal recess means for
instrumentation purpose, generally the system shall be
designed for clamping to internal side wall of spun piles;(ii) In use, the system shall be able to be pneumaticallyactuated and control remotely above the ground and theseries of anchors in access hole shall be able to beconnected by means of interconnecting rods.
(iii)The system shall be designed to be retrievable from theaccess hole, with allowance for eight retractable pistons
per anchor.
(iv)The anchor shall be able to be clamped to the side wall byat least 300 psi pressure.
(v) Materials : Stainless steel or copper.
(vi)The series of anchors in access hole shall have allowancefor connecting to precision transducer sensor by means offittings that grip the interconnecting rods.
The concept of using hollow core of spun piles (or cylinder
concrete piles) as recess means for retrievable
instrumentation approach was then reviewed andre-examined, leading to suggested ideal arrangement of
retrievable instrumentation approach as shown in Figures
3(a),(b) , (c) & (d).

Existing Ground Level
Strain GaugesvLev. A

5

Clay

Instrumented Pipe
Cement GroutSpun Pile

10

Strain Gauges Lev. B
Depth (m)

15

Sandy
Clay

Strain Gauges Lev. C

20

Strain Gauges Lev. D
25

Sandy
Silt
30

SI borehole log

Strain Gauges Lev. E
Strain Gauges Lev. F
Pile toe at 30.0 m depth
Legends:
denotes Vibrating Wire Strain Gauges
denotes Pile Joint

Fig. 2(a) - Diagrammatic illustration of approximate spun pile
instrumentation method

3.1 Description of the State-of-the-Art Global Strain
Extensometer technology

Cement Grout Infill (Usually Grade 25)

The technology consists of a deformation monitoring systemthat uses advanced pneumatically-or hydraulically-anchoredextensometers coupled with high-precision spring-loadedtransducers, and a novel analytical technique to monitor loads
and displacements down the shaft and at the toe of foundation
piles. This method is particularly useful for monitoring pileperformance and optimizing pile foundation design.
To appreciate the innovation contained in the technology, thebasic deformation measurement in the pile by strain gauges
and tell-tale extensometers are reviewed. Normally, strain
gauges (typically short gauge length) are used for strain
measurement at a particular level or spot, while tell-taleextensometers (typically long sleeved rod length) are usedpurely for shortening measurement over an interval (over alength between two levels). From a ‘strain measurement’point of view, the strain gauge gives strain measurement overa very short gauge length while the tell-tale extensometer
gives strain measurement over a very long gauge length!
Tell-tale extensometer that measure strain over a very longgauge length may be viewed as a very large strain gauge orOriginal
Wall
Thickness(usually
Grade 80
Concrete
Instrumented Pipe (Instrumented Cage also
commonly used)
Fig. 2(b) - Section of instrumented spun pile after cement
grout infilling in Approximate Method32simply called Global Strain Extensometer. With recent
advancement in the manufacturing of high-precisionspring-loaded vibrating-wire sensors, it is now possible tomeasure strain deformation over the entire length of piles insegments with ease during static load testing.

Fig. 3(c) Completed improved prototype for application

Fig.3(a)Arrangement of retrievable instrumentation
approach for typical spun pile instrumentation
application.

Fig. 3 (d) Actual Global Strain Extensometer system for spun
pile application
Fig. 4 shows a schematic spun pile instrumentation diagram
using Global Strain Extensometer technology. This system is
equivalent to the conventional method of using 24 no. strain
gauges and 6 no. sleeved rod extensometers, which might not
be possible to be installed satisfactorily due to congestion in
the spun piles.
For the analysis of test data for spun piles using Global Strain
Extensometer technology, the load distribution can becomputed from the measured changes in global strain gauge
readings and pile properties (cross-section area of spun pileand concrete modulus). Load transferred (P ave ) at mid-pointof each anchored interval can be calculated as:
P ave = ε(Ec Ac )(2)
where,
ε = average change in global strain gauge readings;
Ac = cross-sectional area of spun pile section;
Ec = concrete secant modulus in pile section.
With the instrumentation set-up as described in Fig. 4, theFig. 3(b) Design for retrievable pneumatic extensometeranchors and transducer system33

state-of-the-art Global Strain Extensometers system is able tomeasure shortening and strains over an entire section of thetest pile during each loading steps of a typical static pile load
test, thus it integrates the strain over a larger and more
representative sample.
N (blows/30cm)
0

50

100

150

0

systems) and test data acquisition systems for verificationload testing programmes are as shown in Figure 5.

(Pile head)
Existing Ground Level
Anchored Lev. 0
Global Strain Gauge Lev. A
Anchored Lev. 1
Extensometer Lev. 1

5

Clay

Global Strain Gauge Lev. B
Hollow annular space
Spun Pile

10

Anchored Lev. 2
Depth (m)15

Extensometer Lev. 2
Global Strain Gauge Lev. C

Sandy
Clay

Anchored Lev. 3

Extensometer Lev. 3

20

Global Strain Gauge Lev. D
Anchored Lev. 4

Extensometer Lev. 4

Anchored Lev. 5
Anchored Lev. 6

Global Strain Gauge Lev. E
Extensometer Lev. 5
Global Strain Gauge Lev. F
Extensometer Lev. 6

25

Sandy
Silt

30

Pile toe at 30.0 m depth
Legends:

SI borehole log

denotes Glostrext anchored level
denotes Glostrext Sensor
denotes Pile Joint

Fig. 4 - Schematic diagram of typical instrumented spun pileusing Global Strain Extensometer technology
3.2 Verification of the Global Strain Extensometer
SystemFig. 5 Instrumentation Details and Test Setup.
For a typical spun pile section, the test results for Pile Top
Loads Versus Change in Strain measured by both Global
Strain Extensometer Technology and Average of 12 no.
Vibrating Wire Strain Gauges at Levels A, B and C are shownin Figure 6 and , plot of Pile Top Load & Change in Strain
Versus Time for both Global Strain Extensometer
Technology and Average of 12 no. Vibrating Wire Strain
Gauges at Levels A, B and C are shown in Figure 7.
Based on the results obtained from the laboratory tests it canbe concluded that the axial strains measured by the two
independent systems (conventional vibrating wire strain
gauges and Global Strain Extensometers technique) show
very similar characteristics and therefore in good agreement.
The Global Strain Extensometers measure strains over anentire segment of the pile samples, thus it integrates thestrains over a larger and more representative sample than theconventional strain gauges.

For the purpose of verification of the Global Strain
Extensometer system, a special and detailed laboratory
testing programme consisting of a series of full-scale tests on
spun pile sections of various sizes has been implemented. The
tests were carried out and completed in the laboratory,
incorporating Global Strain Extensometer instrumentation
technique alongside with and against well proven vibrating
wire strain gauges systems in the test sections.
The instrumented spun pile sections were tested by thenormal Maintained Load (ML) Method, using hydraulic jackand high strength stress bar system. In the typical set-up used,
the test loads were applied using 1 no. hydraulic jack (1,000
tonnes capacity) acting against the main beam connected byhigh strength stress bar system . The jack was operated by anelectric pump. The applied loads were indicated by calibrated
Vibrating Wire Load Cell. To obtain good quality data, allinstruments readings were logged automatic using Micro-10x
datalogger, at 1 to 2 minutes interval for close monitoring
during loading and unloading steps. The detailed test set-up,
pile instrumentation comparison schemes adopted (using
Global Strain Extensometer technique and strain gauges
34

over a period of approximately two months.
A typical pile instrumentation schemes adopted (using Global
Strain Extensometer technique) for load testing programme,along with nearby borehole SPT N-values plot and subsurface
exploration results from soil investigation are graphically
represented in Figure 8.
Table 1. Test Piles Structural Properties and Installation
Details for Driven Piles

Fig. 6. Plot of Pile Top Loads (kN) Vs Change in Strain (µs)

Fig. 7. Plot of Pile Top Loads (kN) & Change in Strain (µs)
Vs Time (hours) for Spun Pile Section
4.0 FIELD TESTING ON INSTRUMENTED SPUNPILES.
The unique Global Strain Extensometer technology
developed has been used to fully instrument 10 nos. of static
axial compression load tests on driven prestressed spunconcrete pile for a project site in the state of Johor, Malaysia.
The structural properties and installation details (including
drop height and final set record taken) for each test pile are
given in Table 1 below.[8]The high strength prestressed spun concrete piles are driven
with hydraulic impact hammers at the site. The test piles
STP3, STP4R and STP7 were installed with a 9-ton BSP
hydraulic impact hammer, STP5, STP6, STP8R2, STP10R
and STP1 were also installed with a 9-ton BSP hydraulicimpact hammer, while test piles STP2 and STP9R2 were
installed with a 16-ton Twinwood hydraulic impact hammer,

Fig. 8. Pile instrumentation schemes adopted (using Global
Strain Extensometer technique)35

Test results acquired from Global Strain Extensometer
technology on all tested piles appeared to be consistent, andthe typical test results are reproduced here to highlight thecapability of this technique (See Figures 9(a) to (f)). From
highly consistent measurements of the structural elongationof the entire length of piles using Global Strain
Extensometers the pile toe displacement behaviour can bereliably established by subtracting the structural elongationfrom the pile head displacement .
Similarly, Global Strain Extensometer technology also enable
pile displacement under load at any depth along the shaft to beaccurately derived, based on its ability to make segmental or
global shortening/elongation measurement along the whole
pile independent of any external reference and free ofcommon errors associated with the conventional sleeved rod
extensometers. This enables the movement and strains at
various levels down the pile shaft to be accurately
determined, thus permitting a significantly improved means
for the acquisition of load transfer data.
The technology not only enables ease of the assessment ofmodulus–strain relationship and load transfer study it alsosignificantly improves the reliability of the measurement of
movement of pile between deeper soil stratums, as illustrated
in the characteristic curves of mobilized unit shaft friction .

Fig. 9(b). Load Distribution Curve

Fig. 9(c). Load Transfer Characteristics during Loading
stagesFig. 9(a). Plot of pile top load versus pile top settlement, pilebase settlement and total shortening

Fig. 9(d). Plot of Mobilised Unit End Bearing versus Pile
Base Settlement36

methods.
For driven spun piles test results analyzed in this study, theresults of the load transfer parameters for silty sand layer are
summarized in the corresponding correlation of SPT-N
values versus Maximum Mobilised Unit Shaft Resistance fordriven spun piles under this study as shown in Figure 10. The
results of the end bearing parameters for driven spun piles are
summarized in the plot of SPT-N values versus Maximum
Mobilised Unit End Bearing Resistance under this study as
shown in Figure 11

Fig. 9(e). Plot of Applied Pile Top Load, Total Shaft
Resistance and Base Resistance versus Pile Top
Settlement

Fig. 10 Correlation of SPT value versus maximum mobilized
unit shaft friction in silty sand layer for driven spunpiles.

Fig. 9(f). Plot of Pile Base Load over Applied Pile Top Loadversus Pile Top Settlement
4.1 Correlation between SPT N-Value and Pile Design
Parameters for Driven Spun Piles Test Results
With critical evolution in the understanding of the loadtransfer and bearing behaviour of piles in recent years mainly
through analysis of instrumented full-scale load tests, it iscommon that for driven piles, the axial load capacity can beevaluated empirically from correlation with standard
penetration tests SPT N-values (which are widely used in thisregion) using modified Meyerhof Equation, where theultimate bearing capacity of a pile in compression is given byequation (1)As the instrumented full-scale static load tests for driven piles
are rarely being carried out in this region due to lack ofinnovation in this area, there is strong need for the evaluationof Ks and Kb factors for driven piles to be carried out moreroutinely for the improvement of empirical method ofdesigning piles based on the results of full-scale instrumented
load tests, to catch up with the evolution in the design

Fig. 11 Correlation of SPT value versus max. mobilized unit
end bearing in silty sand layer for driven spun piles.
37

The comments on correlation for driven spun piles test results
are as follow:

interpretation tool, capable of leading the spun pile
instrumentation industry to a revolutionary improvement notseen in the past.
Some of the obvious benefits of using Global Strain
Extensometer technology are as follow:a) The silty sand layer for pile founding depth is generally
overlain by 7m to 10m hydraulic sand fill followed bycompressible weak marine clay layer from approximately
10m to 26m depth, and the silty sand layer that underlies
the sand fill and marine clay extends to depths beyond thetoe of all driven spun piles;(i) The technology enables installation of instrumentation
after pile-driving and thus virtually eliminates the risk of
instrument damage during pile production andinstallation;b) The SPT N-value in this silty sand layer generally varies
between 10 and 100. The maximum mobilized unit shaft
friction from the test results is generally between 29 kpa to
346 kpa;(ii) The post-install nature of the method empowers engineers
to select instrumentation levels along the as-built depth offoundation piles using pile driving/installation records
and site investigation data as guides;c) Correlation results for driven spun piles yield shaft
resistance factor, Ks = 3.2 in the shaft resistance empirical
relationship of fs = Ks*Ns, where fs is the ultimate unit
shaft reisistance, Ks is the empirical design factor relating
ultimate shaft load to SPT N-value, kN/m2 per SPT blow
and Ns is the average SPT value for the pile shaft,blows/300mm.

(iii)The
technology reliably
measuressegmental
shortening/elongation and strain over an entire section of
the test pile during each loading step of a typical staticload test. Unlike the conventional strain gauges that makejust localized strain measurements, the new technology
integrates individual measurements over a larger andmore representative sample;d) Correlation results for driven spun piles yield end bearing
resistance factor, Kb = 1.3 in the end bearing resistanceempirical relationship of qb = Kb*(40Nb), where qb is
the ultimate unit end bearing reisistance, Kb is theempirical design factor relating ultimate end bearing load
to SPT N-value, kN/m2 per SPT blow, Nb is the SPT
value for the pile base, blows/300mm.

(iv)Significant cost and time saving, as the additional cage
and cement grout infilling are not required;(v)The technology is extremely environmental friendly, as thesensors are retrievable, and no messing around withcement grouts; and5. CONCLUSIONS.

e) It is worth to note that Singapore Standard (SS) CP4: 2003
code of practice for foundations’ recommendation for Ks
is between 2 to 5 for driven piles shaft resistance factor,
subject to a limit of fs = 200 kpa, and Kb is between 6 to 9
for driven piles end bearing resistance factor, subject to alimit of qb = 18,000 kpa.

The successful driven field testing programmes using theunique Global Strain Extensometer pile instrumentation
technique have clearly demonstrated that the novel Global
Strain Extensometer instrumentation technique provides highquality, reliable and consistent results. Three distinct featuresof this method would especially appeal to engineers:• The method enables installation of instrumentation after
pile-driving and thus virtually eliminates the risk of
instrument damage during pile production and installation.
• The post-install nature of the method enables engineers to
select instrumentation levels along the as-built depth ofdriven piles using pile driving records and site investigation
data as guides.
• The method reliably measures segmental shortening andstrains over an entire section of the test pile during eachloading step of a typical static load test and unlike
conventional strain gauges that make just localized strain
measurements, integrates individual measurements over alarger and more representative sample, thus making the test
results more informative.
Pile instrumentation using the Global Strain Extensometer
technique developed enables acquisition of design data in amore accurate and reliable way for shaft friction and end
bearing parameters to be obtained for driven piles andsubsequently help to optimize pile design. It also helps tof) Normally the critical shaft displacement is relatively
small to fully mobilize the shaft resistance compared tothe big movement that is needed to fully mobilize the end
bearing. Therefore Ks = 3.2 from this study is able toserve as a useful initial design guide for shaft resistancefactor in silty sand materials, while Kb = 1.3 represent a
conservative approach for end bearing resistance factor insilty sand materials, as the majority of the piles are nottested to full failure.
4.2 Advantages of using the Global Strain Extensometer
Technology
Due to the significant difference in the methodology
evolution, from conventional sacrificial cast-in method to a
new retrievable post-install approach, the Global Strain
Extensometer technology has been proven via full-scale load
tests to be a reliable and powerful pile load testing and data38

derive correlation of shaft friction and end bearing parameterswith the standard penetration tests commonly used locally.
5 REFERENCES.
[1] Fleming, K., 1992, “A New Method For Single Pile
Settlement Prediction and Analysis,” Geotechnique 42,
No.3, 411-425.
[2] Fleming, K., 1996, Talking Point: Ken Fleming Assesses
the Present Major Issues in the Pile Testing Industry,Page
3. Ground Engineering, October 1996.
[3] Singapore Standard on Code of Practice for Foundations,
CP4:2003[4] Chan, S.F & Lee, P.C.S., 1990, “The Design ofFoundations for Suntec City, Singapore,” Proceedings ofConference on Deep Foundation Practice, Singapore.
[5] Chan, S.F., 2004, “Special Lecture, Design andConstruction of Foundations for Suntec City, Singapore,”
Proceedings of the Malaysian Geotechnical Conference,
Malaysia, pp 21-43.
[6] Aziz, H.M. A. & Lee, S.K., 2005, Innovation in
Instrumented Test Piles in Malaysia: Application of
Global Strain Extensometer (GLOSTREXT) Method for
Instrumented Bored Piles in Malaysia, Bulletin of theInstitution of Engineers, Malaysia, October 2005 issue. pp
10-19.
[7]Aziz, H.M. A. & Lee, S.K., 2006, “Application of Global
Strain Extensometer (GLOSTREXT) Method for
Instrumented Bored Piles in Malaysia,” Proceedings of
10th International Conference on Piling and Deep
Foundations, Amsterdam, pp 669-767[8] Lee, S.K, 2011, “Development and Application of a New
Instrumention Technique for Spun Concrete Piles” Ph.D
Thesis, University of Malaya.
.

39

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Unique Grouting Material Composed of Calcium Phosphate Compounds
S. Kawasaki1 and M. Akiyama 2
1
Hokkaido University, Japan2
Geoscience Research Laboratory Co., Ltd., Japan

ABSTRACTTo employ calcium phosphate compounds (CPCs) in ground improvement measures such as the reinforcement of soil
and rock, we examined suitable conditions for CPC precipitation and performed unconfined compressive strength(UCS) tests of sand test pieces cemented by CPC. Two types of phosphate stock solution and two types of calcium stock
solution were used to prepare the reaction mixtures and CPC precipitation was confirmed in all reaction mixtures. The
volume of CPC precipitation in the mixture increased as the pH rose from strongly acidic to around neutral. The UCS ofsand test pieces cemented by 1.5 M diammonium phosphate and 0.75 M calcium acetate tended to increase with time,reaching a maximum of 87.6 kPa. The results indicate the practical feasibility of using unique and new CPC grouts as
chemical grouts because of their self-setting property and as biogrouts because of the pH dependence of precipitation.
Keywords: calcium phosphate compound, grout material, unconfined compressive strength, self-setting property
apply biogrout to various soils and rocks, it is very important
to increase the number of mechanisms available for the
precipitation of cement materials.
In engineering as well as in science, it is essential to learn
from nature. In nature, various minerals, such as calcium
carbonate, calcium sulfate, calcium phosphate, calcium
oxalate, silicate, and iron oxide are precipitated by livingorganisms. These biominerals are promising as engineering
materials because they have considerable strength and low
environmental impact. In this study, we carried out afundamental examination on novel grout materials composedof calcium phosphate compounds (CPCs) (Fig. 1). CPCs exist
as phosphate rocks (e.g., fluorapatite) in the naturalenvironment and also as an important inorganic substance(e.g., hydroxyapatite, HA) in living organisms [7]. As shownin Table 1, there are 11 known CPCs with various
calcium-to-phosphate (Ca/P) molar ratios in the ternary
system Ca(OH) 2 –H 3 PO 4 –H 2 O. Research and developmentof materials composed of CPCs are currently in progress,
especially in the fields of medicine and dentistry.

1. INTRODUCTION
Novel grout materials have been developed to reinforce theground and/or to control ground permeability with bacterially
produced cement material [1]–[6]. These grout materials arecalled biogrouts. Three mechanisms of mineral formation
have mainly been considered for biogrouts. The firstmechanism is the precipitation of calcium carbonate by in situ
microorganisms and/or added yeasts [2]. In this process,calcium carbonate is precipitated by the binding of carbonate
ions released from microorganisms and calcium ions from thegrout, which includes calcium and glucose. The second
mechanism was reported by Whiffin et al. [3], who used urea
instead of glucose and ureolytic Sporosarcina pasteurii
instead of yeast and other in situ microorganisms; thedecomposition of urea by S. pasteurii produced carbon
dioxide, which supplied the carbonate ions. In both cases,
additional pH buffers or ammonium ions play the role of pH
adjusters for effective precipitation. The third mechanism isbased on the pH dependence of the extension speed of thesiloxane bond; this mechanism was reported by Terajima etal. [5], who utilized the carbon dioxide produced by yeast to
neutralize the alkaline active silica solution because thesiloxane bond rapidly extends and gelates in the middle rangeof pH.
Soil and rock vary infinitely in their physical and chemical
properties. This fact orients the development of biogrout
along two main directions: one, to develop a highly
general-purpose biogrout, and the other, to develop a
specialized biogrout for a specific type of soil or rock. To

Step 1

In vitro experiment using chemical reaction

Step 2

In vitro experiment using a test piece madeby chemical reaction

Step 3In vitro experiment using chemical reaction
and microbial activity

Step 4

In vitro experiment using a test piece madeby chemical reaction and microbial activity

Step 5

Scale-up of in vitro experiment

Step 6

Practical scale experimentThis study

Fig. 1 Flowchart of the study. The steps carried out in thisstudy are highlighted in gray.
40

0.5
1.0
1.0
1.33
1.5
1.5
1.2-2.2
1.5-1.67
1.67
2.0

Compound

Monocalcium
phosphatemonohydrate
Monocalcium
phosphateanhydrate
Dicalcium
phosphate
dihydrate
Dicalcium
phosphateanhydrate
Octacalcium
phosphate
α-tricalcium
phosphate
β-tricalcium
phosphate
Amorphous
calcium phosphate
Calcium-deficienthydroxyapatite
Hydroxyapatite
Tetracalcium
phosphate

Abbreviation
MCPM
(MCP)

Formula

MCPA
(MCP)

Ca(H 2 PO 4 ) 2

DCPD
(DCP)

CaHPO 4 ·2H 2 O

DCPA
(DCP)

CaHPO 4OCP
α-TCP

Ca 8 (HPO 4 ) 2 (PO 4 ) 4 ·5H 2O
α-Ca 3 (PO 4 ) 2β-TCPβ-Ca 3 (PO 4 ) 2

ACP

Ca x (PO 4 ) y ·nH 2 O

CDHA

Ca 10-x (HPO 4 ) X (PO 4 ) 6-x(OH) 2-x (0 < x < 1)
Ca 10 (PO 4 ) 6 (OH) 2
Ca 4 (PO 4 ) 2 O

HA
TTCP

Ca(H 2 PO 4 ) 2 ·H 2 O

1000

Gel

100

ACP
OCP

10
1

HA

0.1

Time

Fig. 2 Formation, stability, and hydrolysis of calcium
phosphates as a function of phosphate concentration (log(P))
in solutions of amorphous calcium phosphate (ACP) at
neutral pH. OCP, octacalcium phosphate; HA,
hydroxyapatite. The figure is adapted from Tung [8].
1
DCPD

0

Medical CPC paste, however, is extremely expensive andhas high viscosity, which makes it unfeasible for engineering
applications. Therefore, we considered CPC use from anengineering viewpoint and aimed to develop a grout material
that could be precipitated under normal temperature andpressure through microbial activity by using materials thatcan be easily handled. To the best of our knowledge, no
existing grout material makes use of the self-setting
mechanism of CPC alone or employs microbial pH
adjustment activity for CPC precipitation. CPCs have unique
physical and chemical properties. Their numerous advantages
as a grout material include the following:(1) Gel-like or amorphous CPCs change into HA over time
(Fig. 2 [8]). Therefore, CPC hardens after injection into
soil and rock because of the self-setting mechanism.
(2) The solubility of CPCs depends on the pH of thesurrounding environment (Figs. 3 [8] and 4). This makes
it possible to utilize the mechanisms of pH adjustment by
microorganisms, which are used in known biogrout
methods to control CPC precipitation.
(3) Phosphate and calcium stock solutions can be made fromfertilizers, and calcium and phosphate can also be
extracted from the bones of livestock and the shells ofaquatic animals, respectively.
(4) CPCs that precipitate after grout injection are non-toxic.
(5) Unlike concrete, re-excavated muck that consists of soil,rock, and CPC grout is recyclable as agricultural fertilizer.
We begin by focusing on the development of a novel grout
material intended for soil materials (sand, in this study). We
then extend our approach to control the strength andpermeability of rock materials. This work can contribute to

OCP β-TCPα-TCP
TTCP-1log [Cal]

Ca/P
ratio
0.5

Phosphate concentration (mM)

Table 1 Properties of biologically relevant calcium
orthophosphates. The table is adapted from Dorozhkin and
Epple [7].

DCPA

-2
-3
-4

HA

-5
-6

2

0

4

6
pH

8

10

12

Fig. 3 Solubility phase diagrams for the ternary system,
Ca(OH) 2 –H 3 PO 4 –H 2 O, at 25 °C, showing the solubility
isotherms of CaHPO 4 (DCPA), CaHPO 4 ·2H 2 O (DCPD),
α-Ca 3 (PO 4 ) 2(α-TCP),Ca 8 H 2 (PO 4 ) 6 ·5H 2 O (OCP),
β-Ca 3 (PO 4 ) 2(β-TCP), Ca 4 (PO 4 ) 2 O (TTCP), and
Ca 10 (PO 4 ) 6 · (OH) 2 , (HA). The figure is adapted from Tung
[8].

4.6

41

4.8

5.0

5.2

5.4

5.7
pH

6.0

6.2

6.3

6.5

6.6

Fig. 4 Effect of pH on the precipitation volume of CPC.
countermeasures for liquefaction in alluvial plains andreclaimed land, and to measures for preventing the failure ofnatural and/or artificial slopes.
In this study, the most suitable conditions for CPC
precipitation were determined in an in vitro examination byusing phosphate and calcium stock solutions (Step 1 in Fig. 1).
Subsequently, test pieces composed of sand cemented by
CPC were subjected to unconfined compressive strength(UCS) tests and observed by scanning electron microscopy(SEM) (Step 2 in Fig. 1). Our aim was to evaluate thefeasibility of using the unique and novel grout as both achemical and a microbiological grout by exploiting theself-setting property of CPC and the microbial pH adjustment
activity in CPC precipitation, respectively.

the results of the in vitro CPC precipitation test.
A

C

B

D

2. MATERIALS AND METHODS
2.1 CPC Precipitation Test
Reagents with relatively high solubility were chosen forconvenient handling in a practical application.
Monoammonium phosphate (MAP, pH 4.2) anddiammonium phosphate (DAP, pH 8.0) were used as thecomponents of the phosphate stock solution—a type of
agricultural fertilizer that can be prepared easily. For the
calcium stock solution, calcium nitrate (CN) or calcium
acetate (CA) was used. All reagents were of special grade.
The reaction mixtures were prepared by mixing theammonium phosphate (AP) solution and one of the two
calcium solutions at a 1:1 volume ratio (2 mL each);consequently, the final concentrations of the AP and calcium
stock solutions were half the initial concentrations. The initialconcentrations of the AP and CN solutions were 0.2, 1.0, 2.0,
and 3.0 M (the final concentrations were 0.1, 0.5, 1.0, and 1.5
M). A 3.0 M MAP solution could not be prepared because thewater solubility of MAP at 20 °C is 27.2 g/100 g. The
solubility of calcium acetate is 25.8 g/100 g water (20 °C).
Because we could not prepare a 2 M solution of calcium
acetate, the concentrations of the CA stock solution were set
at 0.2, 1.0, and 1.5 M (the final concentrations were 0.1, 0.5,
and 0.75 M). The in vitro precipitation tests were conducted
for the entire set of combinations of AP stock solutions and
CN or CA stock solutions.
Subsequently, the reaction mixtures were left standing at
20 °C for 3 months to monitor their precipitation status; aperiod of 3 months was chosen because changes in this status
can appear with time depending on the properties of the CPC.
To observe the precipitation status, the proportion of the CPC
precipitation volume to the reaction mixture volume (PPV)
was measured during this time (the maximum PPV was 1.0).
Simultaneously, the pH of the reaction mixture was measured
by using a pH Spear, a device designed for measuring the pHof solids and semisolids, including gels.

Fig. 5 Toyoura sand test pieces cemented by CPC. A, mixing
stock solutions and Toyoura sand; B, a test piece covered with
Parafilm; C, a self-standing test piece cemented by CPC; D,
UCS test.
Table 2 Physical characteristics of Toyoura sand.
Soil particle density, ρ s (g/cm3)
Minimum density, ρ dmin (g/cm3)
Maximum density, ρ dmax (g/cm3)
Maximum void ratio, e max
Minimum void ratio, e min
Mean grain size, D 50 (mm)
10% diameter on grain size diagram, D 10 (mm)Fine fraction content, Fc (%)

2.64
1.335±0.005
1.645±0.010
0.973
0.609
0.17
0.11
0

Criterion 1: Set of stock solutions with the minimum
concentration among reaction mixtures that showed a PPV of
1.0 (DAP/CN = 1.0 M:0.5 M, DAP/CA = 1.0 M:0.5 M).
Criterion 2: Set of stock solutions with the maximum PPV
among those with the maximum concentration (DAP/CN =1.5 M:1.5 M, DAP/CA = 1.5 M:0.75 M).
Criterion 3: Set of stock solutions with a concentration thatis closest to average of the concentrations discussed in criteria
1 and 2 (DAP/CN = 1.0 M:1.0 M, DAP/CA = 1.0 M:0.75 M).
Table 2 lists the physical properties of the Toyoura sandused as a soil material in this study. To avoid the destructionof the test pieces during their removal from the mold, theinner wall of the mold container (φ = 5 cm, h = 10 cm) wascovered with a 0.01-cm-thick overhead projector (OHP) sheet.
The volumes of the test pieces made by the mold container

2.2 UCS Test of Sand Test Pieces Cemented by CPC
Reaction mixtures (sets of each stock solution) for the UCS
test (Fig. 5) were selected according to three criteria, based on42and maximum volume of the reaction mixture injected into
the voids between the Toyoura sand particles were 194.59
cm3 and 73.3 mL, respectively. Considering these values,
36.7 mL each of the DAP and calcium stock solutions weremixed, making their final concentrations half of their initialconcentrations. Immediately after the reaction mixture wasprepared, it was uniformly mixed with weighted Toyoura
sand in a stainless-steel ball for 2 min. This mixture wasdivided into quarters, each of which was placed into a moldwith an OHP sheet. The sand in the mold container was
tamped down 30 times by a hand rammer after each of thefour quarters was placed in the mold. Finally, the edge surfaceof the test piece was molded flat and covered with Parafilm M
to avoid desiccation. The test pieces were cured in an airtightcontainer at a high humidity for 1, 7, 14, and 84 days at 20 °C.
The UCS of the test pieces removed from the mold container
after curing was measured at an axial strain rate of 1 %/minby employing a UCS apparatus. For each curing time, two test
pieces were used to perform the UCS test.

3.3 SEM Observation of Sand Test Pieces Cemented by
CPC
Segments of UCS test pieces cemented by the reaction(E) 0.1M CA
1.0
y = 0.75x - 2.16
8.0
r = 0.73

6.0

PPV

PPV

(A) 0.1M CN
1.0
y = 0.58x - 1.30
8.0
r = 0.82

4.0

2.0

2.0

0

1

2

3

4

6

7

8

0

1
2
3
4
5
pH
(F) 0.5M CA
1.0
y = 3.13x - 11.55
8.0
r = 0.96
4.0

2.0

2.0
1

2

3

4

pH

5

6

7

0

8

PPV

4.0

2.0

2.0
1

2

3

4

(D) 1.5M CN
1.0
y = 1.83x - 2.15
8.0
r = 0.89

pH

5

6

7

8

7

8

2

3

4

pH

5

6

7

8

5

6

7

8

6.0

4.0

0

1

(G) 0.75M CA
1.0
y = 2.65x - 7.46
8.0
r = 0.87

6.0

6

6.0

4.0

(C) 1.0M CN
1.0
y = 1.36x - 2.37
8.0
r = 0.93
PPV

5

6.0

0

PPV

pH

PPV

PPV

(B) 0.5M CN
1.0
y = 1.94x - 4.48
8.0
r = 0.91

2.3 SEM Observation of Sand Test Pieces Cemented by
CPC
Segments of UCS test pieces cemented by the reaction
mixture with the maximum PPV value among the sets of AP
and calcium stock solutions with the maximum concentration
were observed by SEM. The segment of the test piece wasallowed to dry naturally at 20 °C for a few days and was thencarbon-coated by a carbon coater. SEM observations werecarried out at an accelerating voltage of 15 kV and at ×2000
magnification. Simultaneously, elemental analyses of test
piece segments were performed by using an energy dispersive
X-ray fluorescence spectrometer (EDX) with SEM.

6.0

4.0

0

1

2

3

4

pH

Sympols: Concentrations of MAP and DAP, MAP 0.1M,,DAP 0.1M, MAP 0.5M,,DAP 0.5M, MAP 1.0M,,DAP 1.0M,DAP 1.5M

6.0
4.0
2.0
0

1

2

3

4

pH

5

6

7

8

Fig. 6 Relationship between the proportion of CPC
precipitation volume of reaction mixture (PPV) and pH for
calcium nitrate (CN) and calcium acetate (CA). The value isrepresented as the proportion of the precipitation volume tothe entire volume of the reaction mixture (4 mL). In vitro
precipitation test was carried out by mixing equal volumes ofammonium phosphate and calcium nitrate or calcium acetate
solutions.

3 RESULTS
3.1 CPC Precipitation Test
CPC precipitation was observed in all reaction mixtures (Fig.
6). The pH of the reaction mixture ranged from 1.6 to 7.5 for
CN and from 4.6 to 7.9 for CA. The pH showed the highest
values among all combinations of AP and CN or CA when
DAP was used as AP. The PPV showed a tendency toincrease with the pH.

mixture with the maximum PPV among the sets of AP andcalcium stock solutions with the maximum concentration
were observed by SEM (Figs. 7 (C) and (F)). In images
captured at ×76 magnification, particles of Toyoura sandwere covered with precipitation in both the CN and CA stock
solutions (Figs. 8 (A) and (D)). In images captured at ×600
magnification, bridges built between particles of Toyoura
sand by precipitation were observed (Figs. 8 (B) and (E)). At
×2000 magnification, plate-like crystals with diameters of~10 μm were observed in the segment cemented by CN stock
solution (Fig. 8 (C)). In contrast, a collection of whisker-like
precipitations was formed in the segment cemented by the CA
stock solution (Fig. 8 (F)). Through elemental analysis, it was
found that the distributions of phosphorous and calcium
overlapped and corresponded to the distribution of
precipitation in the SEM images.

3.2 UCS Test of Sand Test Pieces Cemented by CPC
Toyoura sand test pieces cemented by six reaction mixture
sets were selected based on the results of an in vitro
precipitation test (Fig. 7). The measured UCS in this study
ranged from 10.2 to 87.6 kPa. The maximum value wasmeasured when the DPA/CA ratio was 1.5 M:0.75 M (Fig. 7(F)). The UCS tended to increase with the curing time. Test
pieces with a DAP/CA ratio of 1 M:0.5 M showed a UCS ofover 20 kPa without any tendency to increase with curing
time. The UCS values of the other four reaction mixtures
were below 20 kPa and showed neither increasing nor
decreasing trends as a function of the curing time.

43

concentration of calcium can be achieved in such alkalineconditions [10]. The mechanism proposed by Kawasaki et al.
[2] of utilizing acidification by carbon dioxide from yeasts(A) DAP : CN = 1.0M 0.5M

(D) DAP : CA = 1.0M : 0.5M

100

80

UCS (kPa)

UCS (kPa)

100

60
40
20
0

0

25

50
75
Cure time (day)

25

50
75
Cure time (day)

100

100

80

UCS (kPa)

UCS (kPa)

0(E) DAP : CA = 1.0M : 0.75M

60
40
20

80
60
40
20

0

25

50
75
Cure time (day)

0

100

0

25

50
75
Cure time (day)100(F) DAP : CA = 1.5M : 0.75M(C) DAP : CN = 1.5M : 1.5M

100

100
80

UCS (kPa)

UCS (kPa)

40

0

100

100

60
40

80
60
40
20

20
0

60

20

(B) DAP : CN = 1.0M : 1.0M

0

80

0

25

50
75
Cure time (day)

100

0

0

25

50
75
Cure time (day)100

Fig. 7 Temporal variations in unconfined compressive
strength (UCS) of Toyoura sand test pieces cemented by
CPCs. Mixing concentrations of diammonium phosphate(DAP) and calcium solutions (calcium nitrate (CN) andcalcium acetate (CA)) for mixture of each grout: (A), 1.0
M:0.5 M; (B), 1.0 M:1.0 M; (C), 1.5 M:1.5 M; (D), 1.0 M:0.5
M; (E), 1.0 M:0.75 M; (F), 1.5 M:0.75 M. The dashed line
represents the change in the average of two measurements.

Fig. 8 Scanning electron microscope (SEM) images of
Toyoura sand test pieces cemented by diammonium
phosphate and calcium nitrate (A, B, and C) or calcium
acetate (D, E, and F). A and D, ×76; B and D, ×600; and C andF, ×2000. An accelerating voltage of 15 kV was used for theobservations. Bars represent 500, 50, and 10 μm in A and D,
B and E, and C and F, respectively.

4 DISCUSSION
4.1 Relationship between CPC Precipitation and pH of
Reaction Mixture
The results of an in vitro CPC precipitation test showed that
the PPV tended to increase as the pH increased from strongly
acidic to around neutral. Apparently, this is why the solubilityof the CPC decreased at around neutrality and in a weakly
alkaline region [8]. The Ca/P molar ratios in all reaction
mixtures of each set were constant, regardless of pH. This
indicates the possibility that pH-increasing actions, including
microbial activity, e.g., urea hydrolysis of S. pasteurii [3], caneffectively promote crystallization of CPC after the injectionof an acidic reaction mixture with little precipitation into soil
and rock.
In the case of 1.5 M CN (Fig. 6 (D)), the pH of the reaction
mixture was below 4, even at the maximum PPV. This showsthat if calcium is saturated, CPC can be precipitated in acidic
conditions under which the solubility of CPC is higher. Some
types of peat soils, which typically form a poor groundsurface, show acidic properties due to the presence of humic
acid [9]. The above results demonstrate the suitability ofapplying the CPC grout to such a soil. Because the solubilityof CPC is also higher in alkaline regions around pH 12, a highmay precipitate a large amount of CPC if an alkaline stock
solution is initially used. In this case, it is necessary to study
the utilization and activation of alkali-tolerant yeasts andbacteria.
4.2 UCS Improved by CPC
The most suitable combination of AP and calcium stock
solution concentrations for improving the strength of the UCS
test piece was a DAP/CA ratio of 1.5 M:0.75 M (Fig. 7 (F)),
for which the UCS reached a maximum of 87.6 kPa. In
contrast, all UCS values of the test pieces subjected to CN
treatment were below 20 kPa and showed no significant
trends (Figs. 7 (A)–(C)). SEM images of test pieces subjected
to CA treatment showed whisker-like crystal formation
among particles of Toyoura sand (Fig. 8). It has been reported
that HA whiskers are formed by adding an acetic acidsolution to amorphous calcium phosphate [11]. In Portland
cement, the formation of ettringite, which shows whisker-likecrystals, promotes solidification and increases strength [12].
These results suggest that the strength of the test pieces
subjected to CA treatment in this study might increase if
44

whisker-like HA crystals are formed within them. In addition,
they show that CA would be a potent candidate calcium stock
solution in applications of CPC grout. In contrast, plate-likecrystals were formed in test pieces treated with CN (Fig. 8).
Zhang et al. [13] reported that plate-like HA formed in cases
where CN was used as the calcium stock solution. There are a
few CPCs, e.g., octacalcium phosphate (OCP) [14] anddicalcium phosphate (DCP) [11], whose precipitation leads toplate-like crystals. Non-HA CPC ultimately changes into HA[8]. Thus, the crystal form in test pieces treated with CN maychange after a long curing period, thereby increasing thestrength of the test pieces.
Whiffin et al. [3] reported that the precipitation of at least
60 kg/m3 of calcium carbonate is required to achieve ameasurable strength improvement in a sand test piece(porosity, 37.8%) by using biogrout. Assuming that HA(Ca/PO 4 /OH molar ratio of 5:3:1) precipitates maximally inthe most porous of the Toyoura sand test pieces, thetheoretical value of CPC precipitation in 73.3 mL of thereaction mixture was 5.5 g in the test piece shown in Fig. 7 (F)(porosity, 37.7%). The details of this piece, which iscemented with a DAP/CA ratio of 1.5 M:0.75 M, are as
follows:(1) 0.75 M Ca corresponds to ~2.2 g of Ca mass
(2) 0.45 M PO 4 corresponds to ~3.1 g of PO 4 mass
(3) 0.15 M OH corresponds to ~0.2 g of OH mass
The volume of the Toyoura sand test piece was 194.59
cm3; therefore, the proportionate CPC precipitation mass for a
volume of 1 m3 is 28.3 kg. This means that the CPC grout
attained a UCS of 87.6 kPa by a relatively low amount of
CPC precipitation. A surplus of phosphate may exist in thepore water of the test pieces treated with CA because HA hasa calcium-phosphorus molar ratio of 1.67 (Table 1). Thissuggests that multiple injections of calcium stock solution
increase HA precipitation and induce the strengthening of test
pieces, as described previously [6].
In this study, we begin by focusing on the development ofnovel grout material intended for soil materials (sand, in thisstudy) and assume that the phosphate solution and calcium
solution are mixed just before injection or that they are mixed
in the ground after being sequentially injected. The procedure
for mixing them might be formulated based on the viscosityof the reaction mixture, the PPV of CPC precipitation, and theprecipitation status over time for a particular combination ofstock solutions. It is also necessary to conduct a detailed
evaluation of the relationship between the UCS and the rate ofstiffening because temporal variation in the UCS over thelong term may prove to be one of the most importantparameters in determining the applicability of CPC.

phosphate) and DCP led to a stronger HA with the greatest
purity. The pastes also possessed the properties of self-setting
and biocompatibility.
Research on CPC paste has established that a maximum
compressive strength of 56 MPa can be obtained with amixture of DCP and α-TCP (α-tricalcium phosphate) pastes
and a seed crystal of calcium carbonate; this is substantiallylarger than the compressive strength of 35 MPa obtained
without the seed [16]. This observation shows that thepresence of the calcium carbonate seed crystal can
supplement the effect of strength reinforcement, as is the casefor the CPC grout studied here. Co-existing ions (e.g.,magnesium ions) with in situ pore water change the form of
the CPC crystal [17], and the compressive strength of HA is
increased by the co-existence of bone marrow cells and CPC[18]. These observations suggest that the effective utilizationof ionic and/or organic materials can be an important method
for strength reinforcement in the geotechnical field.
Nowadays, clinical breakthroughs are made daily in the
medical and dental sciences; these include the enhancementof the injectable [19], biocompatibility, and self-setting
properties of CPC paste by organic matter [20] as well as thedevelopment of novel apatite blocks [21]. The injectability of
CPC grout is also one of the most important factors in its
application to natural soils and rocks with minute cracks. The
medical and dental sciences have accrued an extensive
amount of fundamental knowledge regarding CPC.
5 CONCLUSION
In this study, we examined the feasibility of a new grout
material composed of CPC. A candidate cement type that
increased the UCS of the sand test piece with time was
identified by evaluating the effect of pH, the type ofphosphate and calcium stock solutions, and the component
ratios of the phosphate stock solution on the PPV of CPC.
Although only chemical reactions for CPC precipitation were
examined, the results of this study showed that CPCs havesufficient potential for use as (1) chemical grouts, because of
their self-setting property, and (2) biogrouts, because of theircrystal structure and the pH dependence of precipitation.
Changes in the concentration of the reaction mixture were
not reflected proportionally in the strength of the sand test
pieces. In the future, additional tests aimed at determining theimprovement in the strength by CPC are needed to understandmore clearly the underlying mechanical processes and to
facilitate practical application. The relationship between thestrength and the various CPC precipitation parameters(concentration and pH of reaction mixture, curing time, etc.)
should be examined in further detail, as continued research isneeded to identify the process or processes that link crystal
precipitation to the increase in strength. Furthermore,shearing and permeability tests using pieces cemented by
CPC should be conducted to evaluate the applicability of
CPCs for purposes such as permeability control andreinforcement of soil and rock.

4.3 Application of Knowledge in and Collaboration with
Medical and Dental Sciences
The feasibility of improving the UCS was examined byforming CPC with only solutions that can be conveniently
handled during practical application. Chow [15] reported thatthe utilization of CPC seed crystal caused the strengthening of
HA in CPC solution. In particular, Chow [15] pointed out thatutilization of a powder mixture of TTCP (tetracalcium45[14] Wang J, Layrolle P, Stigter M, De Groot K, “Biomimetic
and electrolytic calcium phosphate coatings on titanium
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[15] Chow LC, “Development of self-setting calcium
phosphate cements,” The Centennial Memorial Issue of
the Ceramic Society of Japan, Vol. 99, 1991, pp.
954–964.
[16] Fernández E, Gil FJ, Best SM, Ginebra MP, Driessens
FCM, Planell JA, “Improvement of the mechanicalproperties of new calcium phosphate bone cements in the
CaHPO 4 -α-Ca 2 (PO 4 ) 2 system: compressive strength andmicrostructural development,” Journal of Biomedical
Materials Research, Vol. 41, 1998, pp. 560–567.
[17] Cheng P-T, Pritzker KPH, “Solution Ca/P ratio affects
calcium phosphate crystal phases,” Calcified Tissue
International, Vol. 35, 1983, pp. 596–601.
[18] Vuola J, Taurio R, Göransson H, Asko-sljavaara S,“Compressive strength of calcium carbonate andhydroxyapatite implants after bone-marrow-induced
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[19] Bohner M, Baroud G, “Injectability of calcium
phosphate pastes,” Biomaterials, Vol. 26, 2005, pp.
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[20] Maruyama M, Ito M, “In vitro properties of achitosan-bonded
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Materials Research, Vol. 32, 1996, pp. 527–532.
[21] Tas AC, “Preparation of porous apatite granules from
calcium phosphate cement,” Journal of Materials
Science: Materials in Medicine, Vol. 19, 2008, pp.
2231–2239.

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46

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Shear Stiffness of Unsaturated Soil under Static and Cyclic Loads
C. W. W. Ng, C. Zhou and J. Xu
Department of Civil and Environmental Engineering, the Hong Kong University of Science and Technology, Clear
Water Bay, Kowloon, Hong Kong Special Administrative Region

ABSTRACT
Shear stiffness of soil is one of the most important geotechnical parameters for predicting ground movements anddynamic responses of many earth structures. In this study, a suction-controlled triaxial apparatus is newly modified toinvestigate anisotropic shear stiffness of an unsaturated soil either subjected to static or cyclic loads. This new apparatus
is equipped with three pairs of bender elements and local Hall-effect transducers. Measured results of anisotropic shear
modulus at very small strains, shear modulus degradation curve at small strains and resilient modulus of an unsaturated
soil under both static and cyclic loading-unloading are reported. Semi-empirical state-dependent equations predicting
anisotropic shear modulus and resilient modulus of unsaturated soil are proposed and verified using experimental
results.
Keywords: unsaturated soil, anisotropic stiffness, shear modulus degradation curve, resilient modulus

1. INTRODUCTION
Soils are natural and environmentally friendly materials andthey can be very economical for constructing fill slopes,
landfill covers, pavements and various earth structures
including retaining walls, dams and foundations. Under
proper design, they can be very robust. Accurate
measurements of soil stiffness at very small strains (0.001%
or less) and small strains (between 0.001% and 1%) isessential for correctly predicting ground movements anddynamic responses of many earth structures at working
conditions (Burland, 1989; Atkinson and Sallförs, 1991; Mair,
1993; Ng and Lings, 1995; Brown, 1996; Malone et al., 1997;
Atkinson, 2000; Mancuso et al., 2002; Ng and Menzies, 2007;
Ng and Yung, 2008; Clayton, 2011; Ng and Xu, 2012). Up tonow, small-strain stiffness of saturated soils has been studied
by many researchers (Atkinson and Sallförs, 1991; Atkinson,
2000; Clayton, 2011). It is found that shear stiffness of
saturated soil depends on many factors such as effective stress,
strain level and recent stress history.
In many parts of the world, soils at and near the Earth’s
surface are often unsaturated and they are subjected toseasonal variations of moisture content (i.e., soil suction)..
Previous experimental studies revealed that shear modulus
G 0 at very small strains of an unsaturated soil is significantly
affected by soil suction (Mancuso et al. 2002). This study islimited to the measurement of a single G 0 value from eachspecimen. The degree of stiffness anisotropy of unsaturated
soil is rarely reported, with the exceptions of Ng and Yung
(2008) and Ng et al. (2009). The former investigated thedegree of stiffness anisotropy of recompacted completely
decomposed tuff (CDT) under both saturated and unsaturated

47

conditions, whereas the latter studied the effects ofwetting-drying and stress ratio on anisotropic shear stiffnessof CDT at very small strains. Ng and Xu (2012) explored theeffects of suction history, which refers to current suction ratio
and recent suction history, on both very small strain shear
modulus and shear modulus reduction curve of an unsaturated
soil.
For pavements, subgrade soils are often unsaturated andthey are subjected to seasonal variations of moisture andcyclic traffic loads. Resilient modulus M R , defined by Seed(1962) as the ratio of repeated deviator stress to recoverable
axial strain in a cyclic triaxial test, is commonly used todetermine soil stiffness under cyclic loads. By testing soil
specimens compacted at different water contents, previous
researchers reported that M R of an unsaturated soil issignificantly dependent on water content (Seed et al., 1962;
Jin et al., 1994; Lekarp et al., 2000; Kim and Kim, 2007).
However, soil suction was rarely measured or controlled inprevious resilient modulus tests.
By using a newly modified suction-controlled triaxial
apparatus equipped with three pairs of bender elements andlocal Hall-effect transducers, measurements of anisotropic
shear modulus at very small strains and shear modulus
degradation curve of an unsaturated weathered soil
(completely decomposed tuff, CDT) at small strains aredescribed and discussed in this paper. Moreover, resilient
modulus of CDT under cyclic loads is investigated.
Semi-empirical state-dependent equations are proposed todescribe and predict anisotropic shear modulus and resilient
modulus of the unsaturated CDT. It should be noted that thecontents of this keynote paper are mainly based on Ng andYung (2008), Ng et al. (2009) and Ng and Xu (2012) and Ng
et al. (2012).

2. THEORETICAL CONSIDERATIONSvs (ij ) = G0(ij ) / ρ

2.1 State-dependent anisotropic shear modulus at verysmall strains
Previous researchers (Roesler, 1979; Stokoe et al., 1995;
Jamiolkowski et al., 1995; Bellotti et al., 1996) found that thevelocity of a shear wave propagating in a completely dry or
saturated soil depends on the principal stresses in the plane ofshear (i.e., stresses in the directions of wave propagation andparticle motion), but it is independent of the stress normal tothe plane of shear. Effects of the two principal stresses in theplane of shear on shear wave velocity are found to be similar(Bellotti et al., 1996). In addition to principal stresses, void
ratio also has an effect on shear wave velocity. Based on thesefindings, Ng and Leung (2007) proposed a semi-empiricalequation for the relationship between shear wave velocity andthe state of a saturated soil:−nv s (ij ) = C ij f (e) p r (σ i '×σ j ' )

n/2

(3)

where ρ is bulk density of the medium and G 0(ij) is elastic
shear modulus in the ij plane. If a specimen is homogeneousand cross-anisotropic, shear modulus of an unsaturated soil,
G 0(ij) , can be obtained from Eq.s (2) and (3) and expressed asfollows (Ng and Yung, 2008):
G=
Cij2 F (e) [0( ij )(σ i − ua ) (σ j − ua ) n(u − uw ) 2b×
] [1 + a
]
pr
pr
pr

(4)

where F(e) is void ratio function relating shear modulus tovoid ratio, given by
F (e) =(1)

[ f (e)]2 (Gs + S r e)
ρw
1+ e

(5)

where e is void ratio; G s is specific gravity of soil; ρ w is waterwhere C ij is a constant reflecting the inherent soil properties density and S r is degree of saturation, which is 1 for saturated
in the ij plane, with dimension m/s; f (e) is a void ratio soil.
function relating shear wave velocity to void ratio; p r is areference pressure, taken as 1 kPa for simplicity in this study; 2.2 Degree of stiffness anisotropy
n is an empirical stress exponent; σ i ’ and σ j ’ are effective If the degree of stiffness anisotropy of a soil is expressed inprincipal stresses in the direction of wave propagation and terms of the ratio of shear modulus in the horizontal plane(G 0(hh) ) to that in the vertical plane (G 0(hv) ), by making use ofparticle motion, respectively.
With the development of unsaturated soil mechanics since Eq. (4), degree of stiffness anisotropy of an unsaturated soil
the 1950s, various stress state variables have been proposed can be expressed as follows (Ng and Yung, 2008):
to describe the behaviour of unsaturated soils (Bishop, 1959;
2
Jennings and Burland, 1962; Fredlund and Morgenstern,
G0( hh )  vs ( hh ) Chh2 (σ h − ua ) n[]
1977; Alonso et al., 1990; Wheeler and Sivakumar, 1995; = =(6)
G0( hv )  vs ( hv ) 
Chv2 (σ v − ua )Houlsby, 1997; Gens, 2010). Based on thermodynamic
theory, Houlsby (1997) showed that a complete description of
It can be seen from Eq. (6) that the degree of stiffness
unsaturated soil behaviour requires at least two stress stateanisotropyof an unsaturated soil depends on inherent
variables. Keeping the interpretation of experimental datamaterial
properties,
C 2hh /C 2hv , and the ratio of net normal
rigorous yet simple, this study adopts the most commonlyused two independent stress state variables (Jennings and stresses in the horizontal and vertical directions,
Burland, 1962), namely net normal stress (σ-u a ) and matric (σ h -u a )/(σ v -u a ). However, the degree of stiffness anisotropy
suction (u a -u w ), where σ, u a , u w are total normal stress, pore appears to be independent of matric suction. If a soil is
air pressure and pore water pressure, respectively. Based on subjected to isotropic stress condition, this equation can bethese two independent stress state variables, Ng and Yung further simplified:(2008) proposed the following expression to capture velocity
2of shear waves travelling in unsaturated soil:
G0( hh )  vs ( hh ) 
Chh2
= =

(7)
G0( hv )  vs ( hv ) Chv2
(σ i − ua ) (σ j − ua ) n2
(ua − uw ) b(2)
vs=
Cij f (e) [×] [1 +]( ij )prpr
pr
Equation (7) suggests that under isotropic stress state,degree of stiffness anisotropy expressed in terms of
where b is an empirical exponent reflecting the influence of G 0(hh) /G 0(hv) only depends on inherent anisotropic material
matric suction on shear wave velocity. This equation allows a properties.
smooth transition between unsaturated and saturated states.
When soil is saturated (i.e., u a -u w = 0), Eq. (2) reduces to Eq. 2.3 Resilient modulus under cyclic loads
Many equations have been proposed to predict M R . One of(1).
The shear wave velocity for a homogeneous the most widely used formulations in pavement engineeringcross-anisotropic elastic continuum with a symmetrical is as follows (Uzan, 1985):
vertical axis subjected to principal stresses in the horizontal
and vertical directions is expressed by the following equation
σ + σ 2 + σ 3 b q cyc c(8)
M R = ar ( 1) ()(Mavko et al., 1998):
p
p
r

atm

48

r

atm

where σ 1 , σ 2 and σ 3 are principal total stresses; q cyc is cyclic
shear stress, defined as the amplitude of change in deviator
stress during cyclic loading-unloading; p atm is atmosphericpressure and parameters a r , b r and c r are regression
coefficients. Clearly, Eq. (8) cannot completely describe M Rof an unsaturated soil since it does not consider the effects ofmatric suction on soil behaviour explicitly.
Following its definition, M R can be considered to beequivalent to secant Young’s modulus of a material (Brown etal., 1987). For an isotropic elastic material, Young’s modulus
is linked to shear modulus G by the formulation of E = 2G
(1+ν), where ν is Poisson’s ratio. It may be reasonable toassume that suction affects shear modulus and resilient
modulus in a similar manner qualitatively. Referring to Eq.
(4), it is logical to propose and revise Eq. (8) for M R of an
unsaturated soil as follows:

MR = M0(u − u w k3
p k1 q cyc k 2) () (1 + a)prprpLoad cell

LVDT (Linear variable
differential transformer)

Frictionless loading rod

Top cap
Coarse porous disk
Specimen
Suction probe(9)q cyc k 2u − u w k3
p k1
) (1 +) (1 + a)prprpHall effecttransducer (radial)

Bender element

Hall effecttransducer (axial)

Bender element (vs(vh))

probe (vs(hh), vs(hv))

High air-entry
value disk

Base pedestal

where p is net mean stress, defined as ((σ 1 + σ 2 +σ 3 )/3-u a ).
Parameters k 1 , k 2 and k 3 are regression exponents. The term
M 0 denotes the resilient modulus at the reference stress state
when p-u a =p r , q cyc =p r and u a -u w =0. It empiricallyincorporates the effects of a number of factors such as
microstructure and void ratio on M 0 , which is equivalent to
C ij and F(e) in Eq. (4). The term (p/p r )k1 quantifies theinfluence of net mean stress on M R . The term (q cyc /p r )k2
reflects the variations of M R with cyclic stress. The term (1+(u a -u w )/p)k3 accounts for the influence of matric suction on
MR.
When q cyc approaches 0 (i.e., at very small strains), qk2cyc
approaches infinite because the power k 2 is negative.
Therefore, M R predicted by Eq. (9) approaches infinity whenq cyc approaches 0. This prediction contradicts experimental
observations that soil stiffness at very small strain levels is
finite (Mancuso et al., 2002; Ng and Yung 2008). The
limitation of this equation may be overcome by replacing theterm (q cyc /p r ) by (1+q cyc /p r ). Then, this equation can be
rewritten as:
MR = M0(

Bender element (vs(vh))

Air pressure

Cell pressure

Water outlet

Water inlet

Fig. 1. Modified suction-controlled triaxial apparatus fortesting unsaturated soil (modified from Ng and Yung, 2008)(10)

In this paper, Eq. (10) is proposed to predict
stress-dependent M R of an unsaturated soil.
The validity of the above state-dependent equations, forpredicting anisotropic shear modulus at very small strains (Eq.
(4)), degree of stiffness anisotropy (Eq. (7)) and resilient
modulus (Eq. (10)) of an unsaturated soil, are verified
experimentally later.
3. MODIFIED SUCTION-CONTROLLED TRIAXIALAPPARATUS
Fig. 1 shows the schematic diagram of a modified triaxial
system for testing saturated and unsaturated soils (Ng and
49

Yung, 2008; Ng et al., 2012). The system consists of a triaxial
cell, four independent pressure controllers, five transducers
and a computer. The four pressure controllers are capable ofcontrolling axial stress (σ a ), cell pressure (σ r ), pore airpressure (u a ) and pore water pressure (u w ) independently.
The five transducers consist of an external load cell formeasuring axial force, a linear variable differential
transformer (LVDT) for measuring axial displacement
externally and three pressure transducers for monitoring cellpressure (σ r ), pore air pressure (u a ) and pore water pressure(u w ). This triaxial system is controlled by a closed-loop
feedback scheme. It can be used to apply both static andcyclic loads. To measure water volume change of a soil
specimen, a vertical tube is connected to the pedestal todetermine the amount of water flowing in or out of the soil
specimen. This is achieved by measuring the height of thewater column by a sensitive differential pressure transducer
inside the tube. In addition, this triaxial apparatus is equippedwith a shear-wave velocity measurement system, a set of
local Hall-effect strain transducers and a high capacity
suction probe.
To control matric suction (u a -u w ) during a test, theaxis-translation technique (Hilf, 1956) is adopted to control(u a -u w ) of each specimen by applying u a and u w
independently. u a was controlled through a coarse low
air-entry value (AEV) corundum disk placed on top of a soil
specimen, whereas u w was controlled through a saturated
high AEV (300 kPa) ceramic disk sealed to the base pedestalof the triaxial apparatus. The conventional pedestal of thetriaxial cell was modified to incorporate both a high-entry
value ceramic disk and a bender element. In addition, aspiral-shaped drainage groove of 3 mm wide and 3 mm deep
connected to the water drainage system was carved in thebase pedestal. It serves as a water channel for flushing air
bubbles that may be trapped or accumulated beneath theceramic disk as a result of air diffusion during long periods of
unsaturated soil testing. Diffused air was removed by flushingde-aerated water along the spiral-shaped drainage groove
beneath the ceramic disk once every 24 hours throughout atest.
3.1 Shear wave velocity measurement system
The shear wave velocity measurement system consists ofbender elements, a function generator, a power amplifier, amulti-channel filter, an oscilloscope and a computer. The
arrangement of bender elements on each specimen are shownin Fig. 2(a). Three pairs of bender elements are installed oneach specimen to measure shear wave velocities in different
planes. Thus, very small strain shear moduli in different
planes can be determined. A pair of bender elements wasinserted into the top and bottom platens to measure thevelocity of vertically transmitted shear wave with horizontal
polarization, v s(vh) . A pair of bender element probes wasdeveloped and inserted into the mid-height of each specimen
to measure velocities of horizontally transmitted shear waves
with vertical and horizontal polarization, v s(hv) and v s(hh) . Eachbender element probe consists of two pieces of benderelements, 9 mm in length and 6 mm in width, in a T-shaped
assembly, as shown in Fig. 2(b). More details of the benderelements probe are given by Ng et al. (2004).

Fig. 3. Particle size distribution of completely decomposed
tuff (CDT) (data from Ng and Yung, 2008)
The input signal consists of a single sinusoidal pulse with afrequency of 2 to 15 kHz. The range of frequency was
selected to obtain a clear signal and to minimize the near field
effect (Sanchez-Salinero et al., 1986). The velocity of a shear
wave propagating in soils in the direction of the wave
propagation, i, and particle motion, j, can be calculated fromthe following equation:v s (ij ) =

Li
t ij

(11)

where L i is current travel distance and t ij is travelling time ofthe shear wave propagating over the distance L i . In this study,
L i is determined as the tip-to-tip distance between transmitterand receiver bender elements (Dyvik and Madshus, 1985;
Viggiani and Atkinson, 1995b). t ij is determined bymeasuring time shift between characteristic points of thetransmitted and received signals (Pennington et al., 2001; Lee
and Santamarina, 2005).
3.2 Hall-effect local strain transducers
In order to avoid possible errors in the conventional externalmeasurements of axial strain using a LVDT, such as seating,alignment, bedding and compliance errors (Scholey et al.,1995), Hall-effect transducers (Clayton et al., 1989) are
adopted to measure local strains at center portion of each soil
specimen. As shown in Fig. 1, the triaxial apparatus isequipped with two axial and one radial Hall-effecttransducers for measuring local axial and radial deformationsof a specimen. After calibration by a micrometer, theresolution and accuracy of each Hall-effect transducer isabout 1 and 3 μm, respectively. In the axial strain
measurement, 3 μm corresponds to a stain of about 0.004% inthis study.

Fig. 2. Details of bender element (Ng et al., 2004) (a)Arrangement of bender elements; (b) Details of horizontal
bender element probe at the mid-height of soil specimen

3.3 High capacity suction probe
A high capacity suction probe is developed and used in thisstudy for measuring pore water pressure at the mid-height ofsoil specimen. It is used to provide a guideline for thetermination of suction equalisation stage. When suction
equalisation is reached, u w measured using suction probe isequal to u w applied from the base pedestal. The suction probe
50dimensions of the specimen are measured by a caliper(readable to 0.01 mm) and a PI tape (readable to 0.01 mm),
respectively. The average initial suction of the specimens
after compaction is 95 kPa as measured by a high capacity
suction probe. The variations of measured initial suction
among different height of each specimen and the variations ofmeasured initial suction among different specimens are allless than ±2 kPa.

Table 1. Index properties of CDT (Ng and Yung, 2008)
Index test
Standard compaction tests
Maximum dry density: (kg/m3)
Optimum water content: (%)
Atterberg limits (grain size < 425 µm)
Liquid limit: (%)
Plastic limit: (%)
Plasticity index: (%)

Measured value
1760
16.3
43
29
14

5. TEST PROGRAM AND PROCEDURES

was modified from Druck PDCR 81 miniature pore waterpressure transducer, which mainly consists of a ceramic tip，a small water reservoir and a diaphragm. The original low
air-entry value ceramic tip was replaced by a ceramic tip witha higher air-entry value (500 kPa). It is able to measure anegative pore water pressure of up to 480 kPa, close to the
AEV of the ceramic tip used in the probe, as verified in freeevaporation tests. More details of its development, saturation
procedures and performance can be referred to Ng and Xu(2012).
4. TEST MATERIAL AND SPECIMENPREPARATION
The material investigated in this study is yellowish-browncompletely decomposed coarse ash tuff (CDT) (GeotechnicalEngineering Office, 1988). CDT is weathered saprolite,
commonly found in Hong Kong and often used as
construction materials. Fig. 3 shows the particle size
distribution of CDT as determined by sieve and hydrometer
analyses (British Standards Institution, 1990). The physicalproperties of CDT are summarized in Table 1. Following theUnified Soil Classification System, CDT is classified as silt(ML) (ASTM, 2006).
Fig. 4 shows two stress-dependent soil-water characteristic
curves (SDSWCCs) of CDT. The SDSWCCs were measured
using a triaxial pressure plate system (Ng et al., 2011). Two
identically prepared compacted triaxial specimens were
subjected to a drying and wetting cycle under net meanstresses of 100 and 200 kPa. Total volume change and water
volume changes of each specimen were measured by adouble-cell measurement system and a ballast tube,
respectively. It can be seen from this figure that soil-water
characteristic curve is dependent on stress level. The
specimen subjected to a higher net mean stress tends topossess a higher AEV. The AEVs are estimated to be 60 and
90 kPa at net mean stresses of 100 and 200 kPa, respectively.
In addition, the specimen subjected to net mean stress of 200
kPa shows a slightly lower desorption rate than the one undernet mean stress of 100 kPa. The above observations areprobably because that soil specimen under higher net meanstress has a smaller average pore-size distribution and thus abetter water retention ability (Ng and Pang, 2000).
Each triaxial specimen, 76 mm in diameter and 152 mm inheight, is compacted at initial water content of about 16.3%
and dry density of about 1760 kg/m3. In order to produce a
uniform specimen, the specimen is compacted in 10 layers.
After the completion of compaction, the height and diameter

Four series of tests were carried out on compacted CDT
specimens at unsaturated state. Two of them were conductedto study anisotropic shear modulus at very small strains. The
third series of tests were carried out to study shear modulus
degradation curve at small strains. The purpose of the fourth
series of tests was to study resilient modulus under cyclicloading-unloading.
5.1 Series A tests: isotropic compression test underconstant matric suction
In order to study anisotropic shear modulus of unsaturated
CDT under isotropic stress state and to verify Eq.s (4) and (7),four isotropic compression tests (series A) were carried out atdifferent suctions (Ng and Yung, 2008). The stress path ofeach test is illustrated in Fig. 5. The four specimens (iso-s0,
iso-s50, iso-s100 and iso-s200) were firstly subjected to net
mean stress of 110 kPa and matric suction of 0, 50, 100 and
200 kPa, respectively. Then each specimen was left forsuction equalisation. The purpose of applying a suction
equalisation state is to ensure that soil moisture has reached
its equilibrium state, when pore air pressure and pore waterpressure are equal to predefined values throughout aspecimen. This stage was considered to be completed when
the rate of water content change was less than 0.04% per day(Sivakumar, 1993), which corresponded to a water flow ofabout 0.5 cm3 per day in this study. This equalisation criterion
was used in all series of tests in this study. Once a specimen
was equalised at its predefined stress and suction level, it was
isotropically compressed to each required net mean stress atconstant matric suction. The cell pressure was ramped up at a
rate of 3 kPa per hour to each target value, while maintaining
constant pore air and pore water pressures. This compression
rate was reported by Ng and Chiu (2001) to be slow enough to
AEV=60 kPa
Drying

AEV=90 kPa

Wetting

Fig. 5.
paths of isotropic
compression
tests atcurves
different
4. Stress
Stress-dependentsoil-water
characteristicofmatricsuctions
compacted
CDT
51

ensure fully drained condition by monitoring both changes inwater flow and volume change in a soil specimen at the end ofeach compression stage. Shear wave velocities (v s(vh) , v s(hv)and v s(hh) ) were measured at each predefined net mean stress,
as shown by dots in the figure. More details of this series oftests were given by Ng and Yung (2008).

cyc-s150 and cyc-s250) were dried to matric suctions of 100,
150 and 250 kPa, respectively. Measured resilient moduli
were compared to study suction effects on M R along a drying
path.
During cyclic loading-unloading, net confining pressure
was maintained constant at 30 kPa while applied axial stress

5.2 Series B tests: dying-wetting test under constantisotropic net stress
Two dying-wetting tests (series B) were conducted under two
different constant isotropic net stresses to study the effects ofwetting-drying cycle on anisotropic shear modulus at verysmall strains. Fig. 6 shows the stress paths of the two tests.
Desired net mean stresses of 110 and 300 kPa were first
applied to specimens (dw-p100 and dw-p300), which werethen brought to zero matric suction at constant net mean stressfor equalisation. Shear wave velocities (v s(vh) , v s(hv) and v s(hh) )were measured at the end of the equalisation stage, and thenmatric suction was increased to the next desired value. Uponcompletion of the drying phase at a matric suction of 250 kPa,wetting phase began. When matric suction was decreased tozero, the wetting phase was completed. The drying-wetting
tests consisted of a sequence of equalisation stages. Shearwave velocities (v s(vh) , v s(hv) and v s(hh) ) were measured at theend of each equalisation stage, as shown by dots in the figure.
More details of this series of tests were given by Ng et al.
(2009).

Fig. 6. Stress paths of dying-wetting tests at different
isotropic net stressess (kPa)Measurements of shear
wave velocitiesq300

tc-s300

5.3 Series C tests: suction-controlled constant p triaxial
compression test
Two suction-controlled constant p triaxial compression tests(series C) were carried out on two CDT specimens (tc-s150
and tc-s300) to study effects of matric suction on shear
modulus degradation curve of unsaturated soil. The stresspaths in (p: q: s) space for series C tests are shown in Fig. 7.
Desired net mean stress (100 kPa) and matric suctions (150
and 300 kPa) were first applied to the two soil specimens forequalisation. After the equalisation stage, deviator stress wasincreased at a rate of 3 kPa per hour while the net mean stressand matric suction were kept constant. The shearing rate wasconfirmed by the measurement of the suction probe to be
slow enough that the excess pore water pressure generated
during shearing stage was negligible (less than ±2 kPa). The
use of a constant p stress path was to eliminate the coupling
effects on shear modulus due to a change in the net meanstress (Atkinson and Sallförs, 1991; Ng et al., 2000). Moredetails of this series of tests were given by Ng and Xu (2012).

150

tc-s150

95

0

100

p (kPa)Fig. 7. Stress paths of constant p compression tests at
different matric suctions

5.4 Series D tests: suction-controlled cyclic triaxial test
Six cyclic triaxial tests (series D) were carried out toinvestigate resilient modulus of unsaturated soil under cyclicloads. Fig. 8 shows the stress path of each test in this series.
Three specimens (cyc-s60, cyc-s30 and cyc-s0) were wetted
from initial suction (i.e., 95 kPa) to 60, 30 and 0 kPa at net
mean pressure of 30 kPa, respectively. After suction
equalisation, each specimen was subjected to cyclicloading-unloading to determine M R . Measured resilient
moduli of these three specimens were compared toinvestigate effects of matric suction on M R along a wetting
path. Similarly, the other three specimens (cyc-s100,

Fig. 8. Stress paths of cyclic triaxial tests at different matricsuctions52

was varied with time following a haversine form. For clarity,
variations of axial stress during the first and last 10 cycles areshown in an insert in the figure. According to AASHTO(2003) standard for a resilient modulus test, four levels of
cyclic stress (i.e., 30, 40, 55 and 70 kPa) were considered andapplied to each specimen in succession. At each level ofcyclic stress (q cyc ), 100 cycles of loading-unloading wereapplied at 1 Hz. It should be pointed out that constant watercontent condition was simulated under cyclic loads because
the dissipation rate of excess pore water pressure is low
compared to the rate of repeated traffic loads in the field.
Excess pore water pressure was monitored by a suction probeat mid-plane of each specimen. More details of this series oftests were given by Ng et al. (2012).

al., 2002; Wheeler et al., 2003; Ng and Yung, 2008). This
normal inter-particle

6. INTERPRETATIONS OF EXPERIMENTAL
RESULTS
6.1 Anisotropic shear modulus at very small strains
Fig. 9 shows variations of measured very small strain shear
moduli (G 0(vh) , G 0(hv) and G 0(hh) ) with net mean stress at
different matric suctions (0, 50, 100 and 200 kPa) (obtained
from series A tests). At any suction level, measured shear
moduli, G 0(vh) , G 0(hv) and G 0(hh) , increase with increasing net
mean stress. As given in Eq. (4), shear modulus is
proportional to p2n. As reported by Ng and Yung (2008), n is0.17 for the CDT. With this positive exponent, the form of the
proposed Eq. (4) is consistent with measured data shown in
the figure. Moreover, the rates of increase in shear moduli
with increasing net mean stress are slightly higher whenmatric suction is lower, but the differences become lesspronounced at high suctions ranging from 50 to 200 kPa. For
instances, measured G 0(vh) , G 0(hv) and G 0(hh) at zero suction(saturated state) increase by 144%, 140% and 144%,
respectively, as net mean stress increases from 110 to 400 kPa.
While at a suction of 100 kPa, measured G 0(vh) , G 0(hv) andG 0(hh) increase by 96%, 89% and 94%, respectively, for thesame net mean stress change. Thus, the rate of increase in
shear modulus with increasing net mean stress also depends
on matric suction.
It can be seen from the figure that measured shear moduli,
G 0(vh) , G 0(hv) and G 0(hh) , increase with increasing matricsuction at a given net mean stress. The observed increase in
shear moduli with an increase in matric suction is explainedby the term, [1+(u a -u w )/p r ]2b, in Eq. (4). As reported by Ng
and Yung (2008), the b value is 0.045 for the CDT specimens.
Thus, shear modulus is proportional to [1+(u a -u w )/p r ]0.09.
Similar results that G 0 of unsaturated soil increases withincreasing matric suction have been reported by Cabarkapa etal. (1999) and Mancuso et al. (2000) in their measurements ofsingle G 0 values. The beneficial effects of matric suction onvery small strain shear modulus arise from at least two
possible reasons. Firstly, when a soil specimen becomesunsaturated, voids are partly filled with water and partly
occupied by air, resulting in an air-water interface in eachvoid. When there is an increase in matric suction, the radius ofan air-water interface decreases and hence induces a larger
normal inter-particle contact force (Fish, 1926; Mancuso et53

Fig. 9. Variations of (a) G 0(vh) , (b) G 0(hv) , (c) G 0(hh) with net
mean stress during isotropic compression tests (Ng and Yung,
2008)
contact force provides a stabilizing effect on an unsaturated
soil by inhibiting slippage at particle contacts and enhancing
the shear resistance of the unsaturated soil (Wheeler et al.,2003). Secondly, an increase in matric suction induces thevolume shrinkage of soil specimen (Ng and Pang, 2000). Dueto stronger inter-normal force between particles and higher
density, shear modulus measured at higher matric suctions islarger.
It is also revealed in the figure that measured shear moduli(G 0(vh) , G 0(hv) and G 0(hh) ) increase significantly as matricsuction increases from 0 to 50 kPa at a given net stress.
Beyond a suction of 50 kPa, measured shear moduli continue
to increase but at a reduced rate. Based on the test data given

(v s(vh) ) (Pennington et al., 2001). Thirdly, bender element
embedded in the base pedestal probably deflected more thanthe horizontally mounted bender elements did inside theprobe and hence resulted in a lower v s(vh) (Pennington et al.,2001). Fourthly, platen end effects on stiffness measurements(Germine and Ladd, 1988; Lacasse and Berre, 1988) could be
another cause of the discrepancy. To eliminate the differencesin frequency and boundary effects, G 0(hh) and G 0(hv) reported
in Fig. 9 were chosen for calculating degree of stiffness
anisotropy. This is because that G 0(hh) and G 0(hv) weregenerated by the same bender element probe and they had thesame frequency, travelling path and boundary conditions(Pennington et al., 2001; Ng et al., 2004).
Fig. 10 shows the relationship between degree of stiffness
anisotropy (G 0(hh) /G 0(hv) ) and net mean stress at different
matric suctions (0, 50, 100 and 200 kPa). When matricsuction is zero, G 0(hh) is about 2.5% to 4% higher than G 0(hv)for net mean stress range from 110 to 400 kPa. According to
Eq. (7), this observed stiffness anisotropy should be attributedhv
to the inherent anisotropy (C hh
2 /C 2 ) induced in the tested
material during specimen preparation (Ng and Yung, 2008).
However, this observed stiffness anisotropy may not be solely
caused by the inherent anisotropy induced during specimen
preparation. This is because the specimens have undergone
isotropic compression and suction equalisation, as illustrated
in Fig. 5. The ratio (G 0(hh) /G 0(hv) ) changes during the isotropic
compression and suction equalisation. When matric suction
increases from 0 to 50 kPa, the degree of stiffness anisotropy(G 0(hh) /G 0(hv) ) increases by a ratio of 3.6%, 4.2%, 4.5% and5.1% at 0, 50, 100 and 200 kPa, respectively. On the other
hand, there is only a very small increase (less than 1%) in the
stiffness ratio when matric suction is increased from 50 to 200
kPa. The differences in measured results at different suction
ranges seem to suggest that bulk water effects dominate soil
behaviour when matric suction is lower than AEV of soil
specimen and meniscus water effects dominate soil behaviour
when matric suction above the AEV. Although themagnitudes of increase in G 0(hh) /G 0(hv) are not large, there is aclear trend showing that G 0(hh) /G 0(hv) of CDT increases withan increase in matric suction but at a reducing rate at a given
net mean stress. The measured results seem to be consistent
with theoretical and experimental work by Li (2003) and Mui(2005) who demonstrated that variations of matric suction
could induce anisotropic strains under isotropic stresses.
There is a 1% to 3% increase in G 0(hh) /G 0(hv) when net meanstress increases from 110 to 500 kPa at matric suctions
ranging from 50 to 200 kPa. Although it is generally
recognized that shear modulus will increase with an increasein net mean stress, it is somewhat surprising that the stiffness
ratio also increases with an increase in net mean stress. Thisobserved behaviour may be attributed to the coupling effects
between the hydraulic and mechanical behaviour of
unsaturated soils (Wheeler et al., 2003). As net mean stressincreases at a given matric suction, specific void volume of asoil specimen decreases and water flows out of the specimen
to maintain constant suction. This may provide an
opportunity for re-distribution bulk water and meniscus water
throughout soil specimen. As a consequence of water
redistribution, different mechanical behaviour including

1.10

G0(hh) / G0(hv)

1.08
1.06
1.04
iso-s0iso-s50

1.02

iso-s100
iso-s200

1.00

0

100

200

300

400

500

Net mean stress, p (kPa)

Fig. 10. The influence of net mean stress on G 0(hh) /G 0(hv)during isotropic compression tests (Ng and Yung, 2008)
in Fig. 4, the AEV of the two specimens are estimated to be
60 and 90 kPa at net mean stresses of 110 and 300kPa,
respectively. Soil specimen subjected to a higher net meanstress tends to possess a higher air-entry value. Since the net
mean stresses applied in this series of tests are 110 kPa oreven higher, the AEV of specimens in this series should beequal to or larger than 60 kPa. The observed significant
increase in shear moduli when matric suction increases from0 to 50 kPa (less than AEV of test material) may be because
each specimen remains essentially saturated at any matricsuction less than AEV and so bulk water effects dominate soil
responses (Mancuso et al., 2002). Any increase in matricsuction is practically equivalent to an increase in effective
mean stress, resulting in a smaller void ratio and a much
stiffer soil. On the other hand, as matric suction increases
above the AEV, each soil specimen starts to desaturate,
resulting in the formation of an air-water interface at thecontact points of soil particles. Although the meniscus water
causes an increase in the normal forces which “hold” soil
particles together and hence lead to higher shear moduli, thisbeneficial holding effects could not increase infinitely. This isbecause of progressive degradation in meniscus radius whenmatric suction increases above AEV of a soil specimen(Mancuso et al., 2002; Ng and Yung, 2008).
6.2 Degree of stiffness anisotropy at very small strains
The degree of stiffness anisotropy is expressed in terms of theratio of shear modulus in the horizontal plane (G 0(hh) ) to thatin the vertical plane (G 0(hv) or G 0(vh) ) in this study. In ahomogeneous cross-anisotropic elastic continuum, G 0(vh)should be equal to G 0(hv) . However, a discrepancy between
measured G 0(vh) and G 0(hv) is observed, as illustrated in Figs
9(a) and 9(b). This discrepancy may be attributed to thefollowing four reasons (Ng et al., 2009). Firstly, soil
specimen was actually not a homogeneous cross-anisotropic
elastic continuum. Due to the preferred orientation ofparticles in the horizontal direction, a horizontally transmitted
shear wave travels through fewer particle contacts per unit
distance than does a vertically transmitted shear wave and thisresults in v s(hv) being higher than v s(vh) (Jardine et al., 1999).
Secondly, transmitting frequencies used by the horizontally
mounted bender element probes (v s(hh) and v s(hv) ) were higher
than those used by the platen-mounted bender elements
54

shear stiffness response is expected even if the values of net
mean stress, matric suction and specific volume wereidentical (Wheeler et al., 2003).
Test results shown in Fig. 10 cannot be explained by Eq.
(7) fully. This may be attributed to two reasons: firstly, theproposed equation does not consider any possible coupling
effects between the hydraulic and mechanical characteristics
of unsaturated soils, i.e., possible re-distributions of bulk
water and meniscus water of soil specimens; secondly,
measurements of shear wave velocity are not accurate enoughto differentiate such small degree of stiffness anisotropy. The
estimated measurement error in shear modulus (G 0(ij) ) is
about ±5% (Ng and Yung, 2008).

Fig. 12. The relationship between degree of stiffness
anisotropy and matric suction during drying-wetting tests (Nget al., 2009)

6.3 Effects of drying-wetting on anisotropic shear
modulus at very small strains
Fig. 11 shows variations of G 0(vh) , G 0(hv) and G 0(hh) withmatric suction during two drying-wetting tests (obtained fromseries B tests). It can be seen that G 0(vh) , G 0(hv) and G 0(hh)increase with increasing matric suction following a similar
trend during a drying path. This is consistent with theobservation from series A tests. After drying to the maximum
suction of 250 kPa, soil suction is reduced by wetting the soil
specimen. Similar to SDSWCCs (see Fig. 4), there is ahysteresis between the drying and wetting stiffness curves
showing variations of shear moduli with matric suction. At
any given suction, the shear moduli measured during wetting
are consistently higher than those obtained during drying.
Since the induced axial and radial strains of each soil
specimen are relatively small (less than ±0.3%), total volume
change is not significant. The shear modulus hysteresis may
be because that the water content on the adsorption curve islower than that on the desorption curve at the same suction.
Fig. 12 shows the relationship between degree of stiffness
anisotropy, G 0(hh) /G 0(hv) , and matric suction for the twodrying-wetting tests. At net mean stresses of 110 and 300 kPa,variations of degree of stiffness anisotropy with matricsuction follow a similar trend. Although the magnitudes of
the increase in G 0(hh) /G 0(hv) with an increase in matric suction
are very small, there is a clear trend showing that the degreeof stiffness anisotropy increases with increasing matricsuction. Moreover, G 0(hh) /G 0(hv) value at net mean stress of
300 kPa is consistently higher than that at net mean stress of
110 kPa. These observations are consistent with Fig. 10,

illustrating that stiffness anisotropy is more significant at
higher matric suction and higher net mean stress. There is ahysteresis between the drying and wetting stiffness curves
showing the variations of stiffness anisotropy with matricsuction. The changing rates of G 0(hh) /G 0(hv) and the sizes ofhysteresis loops at two different net mean stresses are almost
the same.
6.4 Shear modulus degradation curve at small strains
The shear modulus degradation curves at small strains arestudied by constant p triaxial compression tests. Fig. 13(a)shows measured stress-strain relationship at matric suctionsof 150 and 300 kPa (obtained from series C tests). Whendeviator stress increases, shear strain of each specimen
increases non-linearly at an increasing rate. Under the samedeviator stress, soil specimen tested at a matric suction of 300
kPa produces consistently smaller shear strain than that at
matric suction of 150 kPa under the same deviator stress. Thisis because the fact that shear stiffness of unsaturated soil
increases with increasing suction level, as illustrated in Fig.s
9 and 11. It is shown in Fig. 13(a) that secant shear modulus(G sec ) of each specimen at small strains can be obtained fromthe slope of a deviator stress-shear strain curve. Theoretically,G 0 may be determined from the initial slope of thestress-strain curve. Due to the limited accuracy of Hall-effecttransducers, G 0 cannot be determined from the stress-strain
curve reliably in this study. Therefore, G 0 measured bybender elements is adopted to represent shear stiffness at verysmall strain level. As illustrated in Fig. 10, measured G 0(hh) isslightly larger than G 0(hv) , meaning that CDT is stiffer in thehorizontal direction. Yimsiri and Soga (2002) proposed a
micromechanics model and demonstrated that the shear
modulus evaluated from triaxial tests is closer to the shear
modulus in the vertical plane (G 0(vh) or G 0(hv) ) when the soil isstiffer in the horizontal direction. Besides, G 0(hv) is believed tobe more reliable than G 0(vh) as discussed in section 6.2.
Therefore, measured G 0(hv) is selected for comparison with
G sec at different shear strains obtained from constant p triaxial
compression tests.
Fig. 13(b) shows the influence of shear strain on G sec in asemi-logarithmic scale at matric suctions of 150 and 300 kPa.
G 0(hv) measured by bender elements before constant pcompression is also shown for comparison. The

Fig. 11. Variations of shear moduli with matric suction
during drying-wetting tests (Ng et al., 2009)55

corresponding shear strain for G 0 is assumed to be 0.001% inthis study (Dyvik and Madshus, 1985). The measured G 0(hv) at
matric suctions of 150 and 300 kPa are 164 and 220 MPa,
respectively. G 0(hv) increases by 34% when matric suction
increases from 150 to 300 kPa. Suction effects on G 0 havebeen explained in section 6.1. Due to the limited accuracy ofHall-effect local strain transducers, strain less than 0.003%
cannot be measured accurately. Thus, there is no reliable datain the strain range from 0.001% to 0.003% shown in thisfigure. When shear strain increases from 0.003% to 1%, G secof the two specimens both reduces significantly but atdifferent rates. The G sec measured at a matric suction of 300
kPa is consistently higher than that obtained at a matricsuction of 150 kPa.

normalized shear modulus degradation curve during constant
p compression tests (Ng and Xu, 2012)
Fig. 13(c) shows secant shear modulus degradation curve
normalized by G 0(hv) . The stiffness could be roughly constant
if the soil behaves elastically as straining proceeds (Simpson,
1992). The shear strain corresponding to the limit of theelastic proportion is generally described as elastic thresholdshear strain (ε e ). Stokoe et al. (1995) reported that ε e of sandis about 0.001% while ε e of clay is much larger (in the orderof 0.005%). In order to avoid subjective assessment, ε e in thisstudy is quantified as the shear strain corresponding to
G sec /G 0(hv) = 0.95 (similar to Clayton and Heymann, 2001).
As illustrated in the figure, the ε e of the specimen tested at amatric suction of 300 kPa is about 0.003%. For the specimen
tested at a matric suction of 150 kPa, although there is no
reliable data shown in the strain range from 0.001% to0.003%, from the trend it can be seen that the ε e is very likely
smaller than 0.003%. The elastic “proportion” of a soil
specimen might represent the proportion of the contacts
between soil particles which are still intact and have notstarted to slide (Simpson, 1992). For soil specimen at a higher
matric suction and lower degree of saturation, more meniscus
water exists in the specimen. Meniscus water causes anincrease in the normal force holding the soil particles together
and resisting the slippage between soil particles. Therefore,
the specimen at a higher suction has a larger ε e . Beyond the ε e ,normalized shear modulus reduces significantly with shear
strain, as shown in the figure. The rate of degradation ishigher for the specimen at higher matric suction. When theshear strain is larger than about 0.02%, the two normalized
shear modulus degradation curves overlap with each other.
This means that the effects of suction on normalized shear
modulus degradation curve become negligible when the shear
strain is larger than 0.02%.
6.5 Effects of number of load applications on resilient
modulus
In series D tests, there are 100 cycles of loading-unloading at
each stress and suction level. To investigate the influence of
number of load applications on resilient modulus, resilient
modulus from the Nth cycle (MNr ) is normalised by resilient
modulus from the first cycle (M1r ).
Fig. 14(a) shows variations of normalised resilient
modulus (MNr /M1r ) with number of load applications (N) after
wetting to zero matric suction (obtained from test cyc-s0 in
Fig. 8). It can be seen that MNr /M1r increases with an increasein N at each cyclic stress (30, 40, 55 and 70 kPa). This is aconsequence of progressive densification resulting from thatthe application of repeated cyclic stress (Dehlen, 1969). In
this study, volume change was determined from axial andradial strains measured using Hall-effect transducers. At zeromatric suction, contractive volumetric strains measured after
100 cycles of loading-unloading are 0.03%, 0.04%, 0.09%
and 0.25%, corresponding to cyclic stresses of 30, 40, 55 and
70 kPa, respectively. The decreasing volume and hence
increasing dry density of each specimen results in an increase
in MNr /M1r with increasing N.
Fig. 14(a) also reveals that the rate of increase in MNr /M1r
with increasing N is dependent on cyclic stress. When cyclic

Fig. 13. Effects of matric suction on (a) stress – strain
relationship; (b) shear modulus degradation curve; (c)56stress is 30 and 40 kPa, M Nr /M 1r increases by about 10%
during the first 20 cycles of loading-unloading. After the first

Fig. 15. Influence of matric suction on resilient modulus
during cyclic triaxial tests (Ng et al., 2012)stress of 70 kPa decreases from 0.25% to 0.03% when matricsuction increases from 0 to 30 kPa. Given such a small
volumetric strain, the variation of MNr /M1r with N becomesinsignificant. At higher matric suction such as 100 kPa, thereis even a slight degradation in MNr /M1r with N. This is becausethat soil dilation rather than contraction occurs under cyclicloads because of suction induced dilatancy (Ng and Zhou,2005). Dilative volumetric strain of -0.03% is measured
during cyclic loading-unloading at cyclic stress of 70 kPa andmatric suction of 100 kPa. Due to the occurrence of dilation,soil density and hence resilient modulus decreases with anincrease in number of load applications. Suction effects on
soil behavior under cyclic loads are further discussed in thenext subsection.
Fig. 14 suggest that the variation of MNr /M1r is negligible
when number of load applications is larger than 20, except
when cyclic stress is larger than 55 kPa at zero matric suction.
An unsaturated CDT specimen generally achieves a stable
resilient modulus within 100 loading-unloading cycles.

Fig. 14. Relationship between normalized resilient modulus
and cyclic number at (a) zero suction; (b) cyclic stress of 70
kPa during cyclic triaxial tests (Ng et al., 2012)
20 cycles, there is no obvious variation of MNr /M1r . For thecase of cyclic stress of 55 and 70 kPa, MNr /M 1r increases
continuously during the 100 cycles of loading-unloading, but
at a decreasing rate after the first 20 cycles. Thus, theinfluence of N on MNr /M1r is more significant at higher cyclic
stress. This is because that accumulated contractive
volumetric strain during cyclic loading-unloading increases
with an increase in cyclic stress at an increasing rate. The
measured contractive volumetric strain is only 0.03% and0.04% at cyclic stresses of 30 and 40 kPa, but increases to
0.09% and 0.25% under cyclic stress of 55 and 70 kPa. Due tolarger contractive volumetric strain, densification effect on
M R is more significant at a higher cyclic stress.
Fig. 14(b) shows the relationship between MNr /M1r and N at
cyclic stress of 70 kPa but different matric suctions (0, 30, 60,
100, 150 and 250 kPa). This figure clearly reveals two typesof soil response. At zero matric suction, MNr /M1r increases
continuously with increasing N. The total increase during the
100 cycles of loading-unloading is up to 20%. On the other
hand, when matric suction is equal to or larger than 30 kPa (s= 30, 60, 100, 150 and 250 kPa), MNr /M1r varies only slightly
with N. One reason is that contractive volumetric strain undercyclic loads is much smaller when matric suction is equal toor larger than 30 kPa. For example, contractive volumetricstrain under cyclic loading-unloading measured at cyclic

6.6 Effects of stress and suction level on resilient modulus
Fig. 15 shows the relationship between M R and s at different
cyclic stresses (30, 40, 55 and 70 kPa) (obtained from series D
tests). M R is the average resilient modulus of the last fivecycles at each stress state (i.e., N = 96-100). Irrespective ofwhether it is along a drying or a wetting path, M R increases
with increasing s significantly. At cyclic stress of 30 kPa, M R
increases by eight times when s increases from 0 to 250 kPa.
The suction effects on soil stiffness have been well illustrated
in Figs 9, 11 and 13(b) and explained in the section 6.1. The
influence of s on M R is captured by the term (1+s/p) k3 in Eq.
(10). Since M R increases with an increase in s, the parameterk 3 should be positive.
Further inspection of this figure reveals that therelationship between M R and s is highly nonlinear along a
wetting path, along which soil suction is smaller than initialsuction. Given the same increase in s, the percentage ofincrease in M R is much larger in the lower suction range. At acyclic stress of 30 kPa, M R doubles when s increases from 0
to 30 kPa, while only increases by 10% when s increases from30 to 60 kPa. Along a drying path, the increase rate of M R
with increasing s is almost constant. The different results
observed in different suction ranges are likely related with57

AEV of a soil specimen, as illustrated in Fig. 10. The different
results observed in different suction ranges are probably
because the bulk water effects dominate soil behaviour whenmatric suction is lower than AEV of soil specimen (hereabout 60 kPa, see Fig. 4) and meniscus water effects dominate
soil behaviour when matric suction exceeds AEV (Ng andYung, 2008).
It is revealed in Fig. 15 that M R decreases significantlywith increasing q cyc at all matric suctions except s = 0. For
instance, M R decreases by about 40 % when q cyc increases
from 30 kPa to 70 kPa at a matric suction of 30 kPa. The
observed decrease of M R with an increase in q cyc is due to thenon-linearity of soil stress-strain relationship (Atkinson,
2000). Fig. 13 shows that soil stiffness is high at very small
strains but it decays with an increase in strain level. In
resilient modulus tests, strain level increases with an increasein q cyc , hence measured M R decreases with an increase in q cyc .
The non-linearity of soil stress-strain behaviour is capturedby the term (1+q cyc /p r ) k2 in Eq. (10). Since M R decreases withincreasing q cyc , the parameter k 2 should be negative. For a soil
specimen with a larger degree of non-linearity, thedegradation of M R with q cyc should be more significant andparameter k 2 should be smaller.

Table 2. Fitted parameters for shear wave propagating indifferent polarization planes (Ng and Yung, 2008)
a

n

b

R2

G 0(hh)

65.5

-0.77

0.17

0.045

0.94

G 0 (hv)

63.2

-0.77

0.17

0.045

0.94

G 0 (vh)

57.7

-0.77

0.17

0.045

0.99

6.7.2 Verification of proposed equations for resilient
modulus
In order to verify the validity of Eq. (10) for representing
resilient modulus of an unsaturated soil, the measured
resilient modulus of four different fine-grained soils are fitted,
including CDT and Keuper Marl, Gault clay and London clay.
M R of CDT is measured in series D tests in this study and thatof the other three soils is reported by Brown et al. (1987).
According to AASHTO classification (2000), Gault clay and
London clay are classified as A-7-5 soil, whereas CDT and
Keuper Marl are classified as A-7-6 soil. Equation (10) was
firstly applied to fit resilient modulus of CDT measured inseries D tests. Four parameters in the equation (M 0 , k 1 , k 2 andk 3 ) are determined by curve fitting using least square method.
This set of parameters except M 0 was applied to fit resilient
modulus of Keuper Marl. The parameter M 0 , which
incorporates the density effect on M R , is determined using theleast square error method. Similarly, Eqn (10) was used to fit
resilient modulus of two A-7-5 soils. The values of soil
parameters except M 0 were determined from resilient
modulus of Gault clay. Then this set of parameters wasapplied to fit measured resilient modulus of London clay. The
basic soil properties and parameter values are summarised inTable 3.
It should be pointed out that this study does not focus on
the influence of net mean stress on resilient modulus. No
cyclic triaxial test is conducted for calibrating k 1 , which is

6.7.1 Verification of proposed state-dependent equationfor very small strain shear modulus
As predicted by Eq. (4), G 0 of an unsaturated soil depends on
void ratio, net normal stress and matric suction. Assumingthat the void ratio function takes the form of a power functionin ea, where a is an empirical void ratio exponent. For
isotropic stress condition, Eq. (4) can be rewritten as
(12)

2b

C ij

modulus can be calculated based on Eq. (12) with the fitted
parameters. The coefficient of determination, R2, for the
calculated and measured shear modulus in each polarization
plane are also summarised in this table. The average value of
R2 is about 0.96, suggesting very strong correlations between
the calculated and measured shear wave velocities. Thus, it isclear that Eq. (12) describes the test results well.

6.7 Verification of proposed equations

G0 (ij ) = C ij2 ρ e 2 a p 2 n [1 + (u a − u w )]Shear wave,
v s(ij)Taking the logarithm of each side gives
ln G0 (ij ) = 2 lnC ij + ln ρ + 2a ln e + 2n ln p + 2b ln[1 + (u a − u w )](13)
Equation (13) is applied to fit very small strain shear moduli
of CDT measured in series A tests (Ng and Yung, 2008). The
parameters C ij , a, n, and b in Eq. (13) can be determined by
the least squares method using a multiple linear regression
model. The fitted parameters are listed in Table 2. Shear

Table 3. Fitted parameters for resilient modulus of four different soils (Ng et al., 2012)Material

AASHTO
classification

Specific
gravity

Plastic
limit

Liquidlimit

Plasticity
index

M0

k1

k2

k3

R2

Se/Sy

CDT

A-7-6

2.73

29

43

14

8.32

1.00

-0.65

1.01

0.98

0.14

Keuper Marl

A-7-6

2.69

18

37

19

6.32

1.00

-0.65

1.01

0.66

0.60

Gault clay

A-7-5

2.69

25

61

36

0

1.00

-0.36

1.31

0.98

0.14

London clay

A-7-5

2.73

23

71

48

0.53

1.00

-0.36

1.31

0.96

0.21

58

and experimental measurements, some key conclusions are
summarised.
From isotropic compression test on compacted soil
specimens at different matric suctions, measured very small
strain shear moduli, G 0(vh) , G 0(hv) and G 0(hh) consistently
increase with an increase in net mean stresses. They also
show significant increases with increasing matric suction at agiven net stress. Moreover, the shear moduli increase moresignificantly below air-entry value (60 kPa). The effects ofbulk water appear to be dominating when matric suction issmaller than the air-entry value whereas the influence ofmeniscus water prevails when matric suction exceeds the
air-entry value.
Measured G 0(hh) , was about 2.5% to 4% higher than G 0(hv)at zero matric suction for net mean stress varying from 110 to
500 kPa. When matric suction was increased from 0 to 50 kPa,
G 0(hh) /G 0(hv) increased by 3.6% to 5.1%. No further significant
increase in suction-induced stiffness anisotropy was observed
at matric suctions higher than the air-entry value. Moreover, asmall increase (less than 3%) in the shear modulus ratio,
G 0(hh) /G 0(hv) was observed when the isotropic net mean stressincreased from 110 to 500 kPa at a constant matric suction
ringing from 0 to 200 kPa. This may be attributed to thecoupling effects of hydraulic and mechanical characteristicsof the unsaturated soil or to sample preparation.
From wetting and drying tests, it was found that there is ahysteresis between the drying and wetting stiffness curves
showing the variations of very small strain shear moduli withmatric suction. At the same suction, the shear moduli
measured during wetting are consistently higher than thoseobtained during drying. The shear modulus hysteresis ispossibly because water content on the adsorption curve islower than that on the desorption curve at the same suction.
There is a clear trend showing that the degree of stiffness
anisotropy increases with an increase in matric suction. Thereis also a hysteresis between the drying and wetting stiffness
curves showing variations of stiffness anisotropy with matricsuction. Although the size of hysteresis loop is very small, thetrend is clear. The changing rates of G 0(hh) /G 0(hv) and the sizesof hysteresis loops at two different net mean stresses (110 and
300 kPa) are almost the same.
From constant p triaxial compression test, shear modulus
of unsaturated soil is consistently larger at higher matricsuction in the shear strain range of 0.001% to 1%. However,
the effects of suction on normalized shear modulus
degradation curve become negligible when shear strain islarger than 0.02%. G 0 , elastic threshold strain (ε e ) and the rateof stiffness degradation all increase with an increase insuction magnitude. This may be attributed to the stabilisation
effect arising from meniscus water when the degree ofsaturation and water content decrease.
For a contractive soil, measured M R of unsaturated soil
increases with the number of load applications under cyclicloads. On the contrary, M R decreases slightly with increasing
number of load applications in dilative soil. An unsaturated
CDT specimen generally achieves a steady resilient response
within 100 cycles of loading-unloading.
Measured M R is highly dependent on stress state. M R
decreases with an increase in cyclic stress due to theFig. 16. Comparison between measured and calculated
resilient modulus of four different soils using newly proposedsemi-empirical equation (Ng et al., 2012)
determined based on previous theoretical and experimental
studies. According to the well-known modified Cam-claymodel, M R (i.e., the axial Young’s modulus during
unloading) is proportional to effective mean stress p' (Muir
Wood, 1990). Viggiani and Atkinson (1995a) also performed
triaxial compression tests to study the relationship between
shear modulus and p'n, where n is regression coefficient. Theyobserved that n is 0.72 at very small strain levels andincreases to 1.0 at a strain level of 0.5%. In this study, thetotal axial strain, including both permanent plastic strain andrecoverable resilient strain, is generally between 0.1% and
1%. Therefore, it may be reasonable to assume that M R
increases linearly with an increase in (p - u a ) (i.e., k 1 = 1) for
simplicity.
Fig. 16 compares the calculated and measured resilient
modulus of these four fine-grained soils. It is well illustrated
that the proposed Eq. (10) is able to capture the variation ofresilient modulus of different soils with stress and suctionlevel. It should be pointed out that two underlying
assumptions are made to keep this equation simple. Firstly,parameters k 1 , k 2 and k 3 are assumed to be constant,independent of stress level and stress history. Secondly,
density effect on M R is incorporated in M 0 . If this equation isused to describe M R of soil specimens with different initial
densities, a single value for M 0 may be not sufficient.
7. SUMMARY AND CONCLUSIONS
Shear stiffness of an unsaturated soil is one of the mostimportant geotechnical parameters for predicting groundmovements and dynamic responses of many earth structures.
In this study, a suction-controlled triaxial apparatus was
newly modified to measure shear stiffness of an unsaturated
weathered soil subjected to both static and cyclic loading
conditions. This apparatus is equipped with three pairs ofbender elements and Hall-effect transducers. Semi-empiricalstate-dependent equations were proposed to capture effects ofnet mean stress, matric suction and void ratio on anisotropic
shear moduli at very small strains and resilient modulus undercyclic loading-unloading. The validity of these equations wasverified by experimental study. Based on derived equations
59

non-linearity of soil stress-strain behaviour, whereas itincreases significantly with an increase in matric suction.
When matric suction is increased from 0 to 250 kPa at a cyclic
stress of 30 kPa, M R increases by up to one order ofmagnitude. This may be attributed to a larger induced
inter-particle normal force resulting from an increase inmatric suction.
8. ACKNOWLEDGEMENTS
The research grant 2012CB719805 of 2012CB719800
provided by the Ministry of Science and Technology of thePeople's Republic of China through the National Basic
Research Program (973 project) is gratefully acknowledged.
In addition, the authors would like to thank the Research
Grants Council of the Hong Kong Special Administrative
Region (HKSAR) for supporting the research grant
HKUST9/CRF/09.
9. REFERENCES
AASHTO 2000. Classification of soil and soil-aggregate.
American Association of State Highway and State
Highway Officials, Washington, D.C.
AASHTO 2003. AASHTO guide for the design of pavement
structures. American Association of State Highway andState Highway Officials, Washington, D.C.
Alonso, E.E., Gens, A. and Josa, A. 1990. A constitutive
model for partly saturated soils. Géotechnique, 40(3):405–430.
ASTM 2006. Standard practice for classification of soils forengineering purpose (Unified Soil Classification System).
American Society of Testing and Materials, WestConshohocken.
Atkinson, J.H. 2000. Non-linear soil stiffness in routine
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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Nanoparticles for Geotechnical Engineering
Mohd Raihan Taha
Dept of Civil & Structural Engineering, Universiti Kebangsaan Malaysia

ABSTRACT
Nanoparticles, one of the results of nanotechnology are very fine materials with wide range of applications. In fields
such as electronics, manufacturing, energy, and medicine its application is well established. However, geotechnicalengineers have not ventured well into nanotechnology although they have understood the importance of nano level
concepts for many-many years. Nevertheless, geotechnical engineers have done research and published on common
nanotechnology topics such as molecular dynamics, nanomechanics, nanoindentation, atomic force microscopy, etc.
Some of these contributions are summarized in this paper. Conventional soil improvement studies using nanoparticles
are also included to show that nanoparticles such as nano-soil, nano-clay, carbon nanotubes, nanoalumina havecharacteristics that can be of advantage to geotechnical engineers.
Keywords: Nanoparticle, nano-soil, geotechnical engineering, soil improvement, geoengineering

1. INTRODUCTION

3.1 Clay Structure

Developments in nanotechnology particularly over the past
10 years have been fully tapped for its advantage by all fields
of knowledge. On the other hand, this cannot be much said forgeotechnical engineering. By its own right, geotechnicalengineering should be at the forefront of nanotechnology as it
is one of the biggest users of materials. In addition,
geotechnical engineers are amongst the earliest“nanotechnologist” in which in the past they have dealt withmaterials at nano sizes and phenomenon that operates at nanoscale [1]. Geoenvironmental engineers are the only group of
geotechnical/soil engineers that are well into applying
nanomaterials in their work, especially for environmental
remediation.
This purpose of this paper is to provide some backgroundon “nano’ related research works in geotechnical engineering.
In addition, some results on the use of nanomaterials such ascarbon nanotubes, nano-soil, nano-clay, nanoalumina, etc forsoil improvement will be covered. As previously mentioned,
geotechnical engineers as one of the major user of materials,
thus should exploit this technology/material to our advantage
and provide efficient use of materials for clients/society.

Based on the fundamental definition of nanomaterial, it is notdifficult to envisage that a clay particle is a nanomaterial.
Furthermore, geotechnical engineers have long recognized
the “nano” dimensions related clay structures. Only
previously, these dimensions were provided in unit ofangstrom (Å) in which 1 Å is 0.1 nm [3].
Molecular issues related to the properties ofnano-meso-microscale montmorillonite clay particulate
systems were studied by Katti et al.[4]. It was shown (firstusing a controlled uniaxial device then electron microscopic
and Fourier transform infrared spectroscopic) that theaddition of solvents causes breakdown of large particles tosmaller nano and meso scale particles with increased water
incorporation in the interlayers causing swelling. In addition,using micro-attenuated total reflectance (ATR) spectroscopy
they found that the increased swelling is related to increased
misorientation of the montmorillonite platelets. Thus, theseinvestigation, observation and behavior can be used todevelop novel nanocomposite systems based onmontmorillonite with tailored properties.
Understanding the nanostructure of clays has lead to thesynthesis of many clay products of various uses. For example,
Chen et al. [5] produced nanostructure nano-clay multi layers
for potential applications such as coatings and membranes
for MEMS/NEMS (micro/nano electrical-mechanicalsystems) and sensors. Nanostructured composite consistingof zirconia and nano-clays is chemically inert, high thermal
resistance, high hardness and the presence of platy nano-claylayer within the structure increases fracture toughness bydeflecting cracks, rendering even higher hardness ofthemultilayers. In geotechnical and geoenvironmental
engineering, these properties may be useful for liners andcaps of nuclear waste containment systems. Thus, with the

2. REVIEW OF RELATED “NANO” STUDIES INGEOTECHNICAL ENGINEERING
Many phenomena that are part of basic geotechnicalknowledge such as soil moisture, water moleculecharacteristics, diffuse double layer systems, soil mineralogy
are processes and materials that operates at nano level [1] &[2]. The following provides some other examples of researchworks in order to show geotechnical engineers involvementin nano related studies.

63

knowledge now available, it is possible to produce any
nanomaterial to suits our needs.

3.3 Nanomechanics
Nanomechanics (nanoscale mechanics) also include
molecular dynamics and its applications in geotechnicalengineering were specifically mentioned by Voyiadjis andSong [13]. It is based on quantum and molecular mechanics
and thus it deals with the equilibrium states of atomic andmolecular forces. Bridging nano-, micro- and macro scales
can be achieved by combination of molecular mechanics,
particular mechanics and continuum mechanics based on thehierarchical approach [14].
However, there are some advantages and disadvantages ofusing nanomechanics in geotechnical engineering.
Advantages include input parameters physical quantities,
such as electron charge, dihedral angle, electron permittivity,
etc. are known and well established [13]. Thus,
nanomechanics provide a new insight into soil behavior
particularly at atomic level. Amongst the disadvantages is theamount of computational time required to arrive at the overallsolution. However, recent developments have reduced thistime to a reasonable period especially with the aid of many
up-to-date softwares now available in the market. Many oftheories are also unfamiliar to many geotechnical engineers.
Therefore, most geotechnical engineers will not have themotivation or capacity to learn and apply them.
Nanoindentation and atomic force microscope (AFM) are
the two main equipment which have found much use in the
study of nanomaterials. Developed in the early 70’s,
nanoindentation was initially designed for testing of small
volume materials. It is simply a device with nano size
indenter providing stress-strain behavior at relatively minuteloadings and displacements. Two main properties as a result
nanoindentation testing are Young’s modulus and hardness.
It is now commonly equipped with a microscope so that theload target area/aim is known. It can also provide “signature”
responses of materials even for materials with very similar
crystal structures [15].
For geotechnical applications, Zhang et al. [16] wrote that
understanding the macroscopic behavior of geomaterials
requires an insight into the micro/nano scale behavior ofindividual particles. Advances in understanding the Young’s
modulus and hardness of individual soil particles cancontribute significantly to micro/nano mechanics andmultiscale modeling of geomaterials.
A schematic diagram of the nanoindentation processes
and load- displacement curve is shown in Fig. 1. For the load
displacement curve, in conventional loading apparatus forsoils (and other materials), the resulting curve is different inwhich load reduces with increasing displacement. In addition,
the resulting plot shows that the behavior of materials is
non-linear even at very-very small loadings.
The AFM is mainly used to study the surface morphology
of materials. In addition, it can be used to measure friction
and almost all surface forces. Using a tip (a mechanicalprobe) which is made of diamond, or silicon or CNT, etc. theprobe moves up and down according to the nature of thesurface during scanning (Fig. 2).

3.2 Molecular Dynamics Simulation
Molecular dynamics (MD) refers to numerical methods ofsolving classical equations of motion (particle dynamics) for
multiparticle systems with known interatomic potentials [7].
Further computer processing and visualization simulations ofresults are commonly known as molecular dynamics
simulations. This approach is particularly popular amongst
nanoscientist in their study of particle interactions in
nanosystems.
In general, the classical Newton’s Law for motion ofparticles is given by

Fi = mid 2 ridt 2(1)where F i is the force acting on particle i, m is the mass of theparticle, r is the distance of the particle i from a datum, t is
time. The force acting on the particle calculated from theinteratomic potential function U ij , may be written as

Fi = ∑ Fij ,i≠ j

Fij = −∇U ij(2)

The potential function can be obtained classically using
Hooke’s Law. However, in MD simulations, nanoscientist
commonly used the Lennard-Jones (L-J) potential function
which gives the total van der Waals force and written as
follows [6] σ 12 σ 6 U ( x) = 4ε LJ −
 r 12 r 6 



(3)

In Eq. (3), ε LJ is the dislocation energy, and σ is the collision
diameter, i.e. the distance in which U = 0. Another common
model involving short-range pair-wise interaction is theMorse potential. A host of other potentials may also be used .
For example, a general 3-D interatomic function also
involving the long range Coulomb, van der Waals interaction,etc. was described by Gutierrez [7]Skipper [8], used MD simulations of clay nano-structure to
study the influence of pore-fluid interaction. The interlayer
region of the stacked clay nano-structure was looked into toprovide insight into the molecular mechanisms underlying
clay swelling.
The influence of applied load on the interlayer ofpyrophillite was studied by Katti et al.[9]. Using steered MD
simulations (SMD), where external forces were applied toindividual atoms, they found that deformation largely
occurred occurred in the interlayer with the clay layer show
relatively little deformation. In another studies [10], [11], &[12], Katti and his co-workers used SMD to study theload–deformation behavior of the interlayer water and natureof water in the interlayer.
64presented and discussed.
4.1 Nano-soil
The nano-soil was obtained by grinding a local soil
(metasedimentary residual soil within the campus ofUniversiti Kebangsaan Malaysia, UKM). After the disturbed
soil was dug and taken to the laboratory, it was first cleared of
all roots and stones, dried in an oven, and sieved. Soil samples
passing through 425 mm sieve was then pulverized in aplanetary mono mill. In this mill, 20 small diameter (5 mmdiameter) balls made of sintered corundum move is a fast
planetary/cyclonic fashion to pulverize the soil. The time ofpulverization was about 10-13 hours in which during thisperiod, distilled water was added to provide over heating.
SEM images of the soil sample before and after milling
showed that there was much more soil in the nano range (i.e.
50-60 nm) in the soils after milling[20]. The specific surfaceof the soil obtained from EGME (ethylene glycol monoethyl
ether) method improve from 2.4 m2/g to 3.9 m2/g. Milling ofkaolinite and montmorillonite samples show similar
observation.
In order to evaluate its potential for improving soil
properties, a soil-cement mix was prepared and the nano-soil
was added to the mix. The amount of nano-soil added was 1
and 2 & by weight of the UKM soil. Thus, the overall mixtureis 94% UKM original soil, 6 % Portland cement, andnano-soil.
The results of compression test on 50 mm3 samples areshown in Fig. 3 [21]. It shows significant improvement in thestrength of the nano-soil mixtures over the control test albeit
the amount of nano soil added was rather quite low. The
strength of the soil-cement mixture almost doubled when 1 %UKM nano-soil was added compared to samples withoutnano-soil. In addition, the mixture which is usually termed as
cement-modified soil in geotechnical engineering will havelower tendency of volume change and PI in addition toincrease in load bearing capacity [21]. This study havedemonstrated that even a small addition of nanoparticles willshow marked enhancement in soil behavior. It is also possible
to engineer nanoparticles to improve behavior of soils to suitdesign and practice requirements.

Figure 1: Nanoindentation processes and typical load-displacement curve [16].
A 3-D image may also be obtained through this imaging
process which can reach down to sub-nanometer range, whichis smaller than the capacity of SEM (scanning electron
microscope). Because it uses a probe, it is possible for theuser to manipulate atoms on materials surfaces. In one
investigation on the pore structure of wet clay [17], there wasa linear decay of pore structure when the water content wasincreased. This was postulated due to the reduction in the
diffuse double layer and more extensive hydrogen bonds
between clay particles and bipolar water molecules.
Microfabric of clay was studied quantitatively using AFM by
Sachan [18]. The quantification was done by developing thecriteria for average and maximum values of angle of particle
orientation using empirical equations for intermediate andextreme microfabrics. Dispersed or flocculated clay
structures were able to be identified for kaolin with varying
content of dispersing agent.

4.1 Carbon Nanotube (CNT)
Carbon nanotube is considered as the first generation
nanomaterial in that it is one of the earliest nanoparticles thatwere produced. Carbon nanotube is so popular and interesting
because it has many “super” properties, especially for thoseworking in electronics and superconducting industries. It has
high unparalleled strength, high electrical and thermal
conductivities, making it ideal for wide range applications[22] & [23]. Thus, it is sometime termed as the “king of
nanomaterial” and also “a solution waiting for problem”. In
order to study the behavior of soil-CNT mix, kaolinite wasused as the soil material. The CNT is a multi-walled nanotube(MWNT) and was purchased from Arkema, France under
product name Graphistrength C100.

Figure 2: AFM scanning probe and its interaction with asurface [19].
4 SOIL IMPROVEMENT WITH NANOPARTICLESIn this section, some results on the applications ofnanomaterials for improving the properties of soil will be
65

Water Content

Liquid Limit

Water Content% CNT

Figure 3: Compressive strength of soil-cement and nano-soilmix.
The black powder CNT material used in this study have abulk density ranging between 50-400 kg/m3. The results ofAtterberg limit tests are shown in Fig. 4 [25]. It is shown thateven with small amount of CNT, the changes in basic
characteristics are measurable. Increase in CNT led toincrease in LL, PL and plasticity index (PI=LL-PL). In thesetests, the addition of the fine CNT particles certainly
increases the overall specific surface of the mixture. As aresult, the need for water, is greater and thus this is indirectly
shown by an increase in LL.
The increase in PI is due to the fact that the increase in PL
is not as dramatic as the increase in LL. It was also found thatthe values of G s , in general, also decrease upon addition of
CNT due to the low density of CNT.
These results have important significance in geotechnicalengineering. In terms of mechanical properties, increase in
LL (and increase in surface area) means the soil will havehigher capacity for water thus decreasing its strength andincreasing its consolidation/compressibility (settlement).
However, increase in PI will generally mean a decrease in its
hydraulic conductivity. For geotechnical structures (such as
foundations and retaining walls) in which strength andcompressibility are the main criteria, the use of CNT may not
be advantageous. In the case of landfill liners and caps, inwhich low hydraulic conductivity is very important, its use
may thus be considered.

Plastic limit

Water Content%

Plasticity Index

Figure 4: Effect of CNT on the Atterberg Limits%ofCNT
kaolinite.
It was also found that the increase in nano-clay gives adecrease in hydraulic conductivity (Fig. 5) due to increase in
the smaller size of particles. The nano-clay particles probably
diffuse between the large sedimentary soil particles, closing
or minimizing the pores which lead to a decrease in the flowof water in the soil mixtures. This is a very significant results
because one of the common materials tested to reduce theshrinkage strain is fiber. Mixing the soil with fiber can reduce
the shrinkage strain up to 90% but the fiber cause an increasein the hydraulic conductivity up to 50 times [26].

4.2 Nano-clay
The performance of nanoparticle in arresting volume changeof compacted residual soil (UKM metasedimentary residualsoil) mixed with a natural montmorillonite was studied using
nano-clay. The nano-clay material was supplied by Southern
Clay Products Inc, Gonzales, Texas, USA under the productname Cloisite®Na+. The size of the particle is 5-15 nm inthickness and 20 nm to 10 µm in diameter.
It was observed that the increase in nano-clay causes small
increase in the plasticity index and linear shrinkage of thesoil. This happened because the nano-clay material is an
expansive clay.

4.3 Nanoalumina
The tests with nanoalumina follow that of nano-clay as
discussed in the previous section. The soil used is the UKM
residual soil mixed with various percentages ofmontmorillonite to produce soil mix with different plasticity
index. Nanoalumina was then added to these soil mixes andits plasticity, shrinkage and hydraulic conductivity wasstudied.

66

Apparently with regard to total volumetric strain, it was
found that that minimum volume change occurs between 0.05
% to 0.1 % nanoalumina content. For high plasticity soils, i.e.
S3 and S4, the crack intensity factor (CIF [26]), is minimum
at 0.1 % (Fig. 7). Furthermore, the hydraulic conductivity
measurements showed that adding nanoalumina will cause aslight decrease in its value (Table 2). Thus, using
nanoalumina for dessication crack control does not cause
detrimental effects on the hydraulic behavior of the soil.
Figure 5: Influence of nano- clay on the hydraulicconductivity
The nanoalumina used was purchased from Infamat
Advanced Materials, Manchester, CT, USA. It has a chemical
formula Al 2 O 3 , particle density of 3.6 g/cm3, surface area of> 150 m2/g, and a transitional gamma crystal structure [27].
The physical properties of the soil mix are shown in Table
1. Adding nanoalumina within 0.1-0.3% (by weight) causes aslight decrease in the plasticity index of the soil mix (Fig. 6).
Maintaining this basic property is important in soil
improvement process as the enhancement of other more
important properties are sought.
Figure 7: Effect of nanoalumina on crack intensity factor.

Table 1: Physical characteristics of the soil mix.
CharacteristicsGs
PI (%)
Passing No. 200
sieve (%)
Clay % (< 2 µm)

Values and descriptions
S2(5) S3(15)S4(20)
2.609
2.612
2.617
28.26
36.18
69.01
49.80
52.44
57.72
47.16

S1 (0)*
2.607
16.96
18

22.9

28.5

Table 2: Hydraulic conductivity values for various
nanoalumina mix.
Soil typeNanoalumina
Hydrauliccontent (%)
conductivity
(m/s)
S1
0
1.86E-09
S1
0.05
1.81E-09
S2
0
1.24E-11S2
0.075
1.30E-11S3
0
1.00E-11S3
0.1
5.40E-12S40
4.56E-12S4
0.1
1.59E-12

38.5

USCS
CL
CH
CH
CL
classification
Compaction
properties
w opt (%)
14.29 16.01
18.63
20.3
18.05 17.70
16.93
16.12
γ max (kN/m3)*The number in the parenthesis indicate the percentage ofmontmorillonite.

5 CONCLUSION
Some general nano level concepts, contributions ofgeotechnical engineers in nanotechnology and use of
nanomaterials for soil improvement are described in thispaper. Concepts such as moisture around a soil particle,
dipolar characteristic of water molecule, diffuse double layer,
and clay structure are some examples of phenomena that
operate at nano level. They have been known and researchedby geotechnical engineers for quite sometime. These serve toshow that geotechnical engineers have long been making
contributions in the field of nanotechnology. The applicationof nanomaterials described in this paper is mainly used forsoil improvement. For example, a small amount of nano-soil
when added to soil-cement mix causes considerable increase
in its compressive strength. The addition of nano-clay and

Figure 6: Effect of nanoalumina on plasticity index.

67

nanoalumina for dessication crack control of soils results insoils with minimal change in plasticity and reduced hydraulicconductivity. Thus, nanomaterials could offer varieties of soil
behavior which may be advantageous for many geotechnicalapplications.

[13] Voyiadjis, G. Z. & Song, C.R., Coupled Theory of
Mixtures in Geomechanics with Application,
Berlin:Springer, 2006, 438p.
[14] Song, C.R., Cho, H., Jung, Y.H., Cheng, A.H.D.,& Alostaz, A., Bridging molecular, particulate, andcontinuum
mechanics
forgeomechanics
application, Advances in Measurement and Modeling ofSoil Behavior (GSP 173), Reston:ASCE, 2007, pp.
[15] Varughese, S., Kiran, M.S.R.N., Solanko, K.A., Bond,
A.D., Ramamurty, U., & Desiraju, G.R., Interaction
anisotropy and shear instability of aspirin polymorphs
established by nanoindentation, Chem. Sci., 2011,2,
2236-2242[16] Zhang, G., Wei, Z., & Ferrel, R.E., Elastic modulus andhardness of muscovite and rectorite determined bynanoindentation, Applied Clay Science, 4, 2009,
pp.271-281.
[17] Selvam, A., See, C.H., Barkdoll, B., Prasad, S., &O’Haver, J., Use of atomic force microscopy for
examining wet clay, Clays and Clay minerals, Vol. 54,
No.1, 2006, pp. 25-28.
[18] Sachan, A., Use of atomic force microscopy (AFM) for
microfacbric study of cohesive soils, Journal of
Microscopy, Vol. 232, Pt 3, 2008, pp 422-431.
[19] Nguyen, C.V., Applications in scanning probe
microscopy, in Carbon Nanotubes – Science andApplications, Meyyappan, M. eds, Boca Raton:CRC
Press, 2005, pp. 137-161.
[20] Taha, M.R., Lim, S.Y., & Chik, Z., Ciri-ciri asasbeberapa tanah yang dikisar (Basic properties of milledsoils), Jurnal Kejuruteraan, UKM, Vol 21, 2009, pp.1-10.
[21] Taha, M.R., Geotechnical properties of soil-ball milled
soil mixtures, Nanotechnology in Construction, Proc
NICOM3, Berlin Heidelberg:Springer-Verlag, 2009, pp.
377-382.
[22] Ozin, G.A., & Arsenault, A.C., Nanochemistry: A
Chemical Approach to Nanomaterials, Cambridge,
UK:RSC Publishing, 2005, 628 p.
[23] Marx, W., & Barth, A., Carbon n nanotubes-A
scientometric study, in Carbon Nanotube, Marulanda,
J.M. eds, Intech Publishers, 2010, pp. 1-18.
[24] Taha, M.R., & Ting, Y., Effect of carbon nanotube on
kaolinite: Basic geotechnical behavior, The Eighteenth
Annual Intl Conf. on COMPOSITES/ NANO
ENGINEERING (ICCE - 18) Anchorage Alaska, USA,Int Community for Composites Enginering, 2010, pp.
721-722.
[25] Taha, O.M.E., & Taha, M.R. Experimental study on the
expansive-shrinkage behavior of residual soil mix with
nano-clay, Proc 5th Jordanian International Civil
Engineering Conference, Jordan Engineers Association,
2012, pp. 417-423.
[26] Taha, M.R., & Taha, O.M.E., Hydraulic conductivity anddessication
crack
behaviorofcompacted
soil-nanolaumina mixture, in NICOM4-Nanotechnlogy
in Construction-4 Intl. Symposium, Konsta-Gdoutos,
M.S., ed., ACI, 2012, 9 p.

6 REFERENCES[1] Taha, M.R., Ismail, E., & Chik, Z., Some nano aspects
and concepts in geotechnology, 2nd Intl Symp on
Nanotechnology in Construction, 2005, pp. 373-381.
[2] Taha, M.R., Nanotechnology in Geoenvironmental
Engineering, Sabbatical Report, Universiti Kebangsaan
Malaysia, 2007, 33 p.
[3] Grim, R.E., Clay Mineralogy, 2nd Ed., New
York:McGraw-Hill, 1968, 600 p.
[4] Katti, D.R., Katti, K.S., & Shanmugasundaram, V., Roleof clay-solvent inter and intraparticle interactions on
swelling
characteristics
ofmontmorillonite
nano-meso-micro scale particulate systems, MRS
Proceedings, 2001, 704:W9.14.1, 6 p.
[5] Chen, H., Zhang, G., Richardson, K., & Luo, J.,
Synthesisofnanostructured
nanoclay-zirconia
multilayers: A feasibility study, Journal of
Nanomaterials, 2008, 8 p.
[6] Liu, W.K., Karpov, E.G., & Park, H.S., Nano Mechanics
and Materials: Theory, Multiscale Methods andApplications, Chichester, UK: John Wiley & Sons Ltd,
2006, 320 p.
[7] Gutierrez, M., Nanogeomechanics:Potential applicationsof nano-mechanics in geotechnical engineering,International Workshop on Micro-Geomechanics AcrossMultiple Strain Scales, Cambridge, UK, March 20-23,
2005, 12 p.
[8] Skipper, N.T., Influence of pore fluid composition on
clay behavior: Molecular dynamics simulations of
nano-structure, in Chemo-Mechanical Coulping in
Clays:From Nano-Scale to Engineering Applications,
DiMaio, C.D., Huckel, T., & Loret, B. (eds)., 2002,
Lisse: Swets & Zeitlinger BV, 2002, pp.65-73.
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Modeling response of pyrophyllite clay interlayer toapplied stress using steered molecular dynamics, Claysand Clay Minerals, 52(2), 2005, pp.171-178.
[10] Schmidt, S., Katti, D., Ghosh, P., & Katti, K., Evolutionof mechanical response of sodium montmorillonite
interlayer with increasing hydration, Langmuir, 21,
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Steered molecular dynamics simulations of dry andhydrated sodium montmorillonite interlayer, Canadian
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pp.836-841.

68

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Damage Assessment and Strengthening of R/C Building Constructed on
Expansive Soils
Osama M. A. Daoud
Building and Road Research Institute, University of Khartoum, Sudan
General Manager of ESCAD Consultancy Co. Ltd.
ABSTRACTThis paper presents an experience in assessment and strengthening of a R/C building carried out in Sudan during the lastyear. A four-story R/C framed hospital building was constructed five years ago in the middle of Sudan, by the White
Nile River. Due to differential upheaval movement of the underneath expansive clay soils (15-25cm), the building had
experienced serious instability problems and severe structural cracks. A comprehensive assessment of the building
skeleton including geotechnical investigation, DT and NDT methods revealed the need of five different strengthening
and retrofitting techniques; there are, transformation of foundation system from isolated footings to strap foundation toincrease the stiffness and rigidity of footings, concrete jacketing of basement floor columns, metal jacketing of thesecond floor columns, construction of shear walls to enhance the stability of the building and finally CFRP laminates
for strengthening of basement floor cover slab. The building had been monitored during and after completion ofstrengthening and retrofitting works to measure any movements during the last year rainy season. Measurements
revealed the efficiency of implemented strengthening and retrofitting techniques.
1. INTRODUCTION

2. LITERATURE REVIEW

Kosti Military Hospital is a four-story R/C framed building
constructed five years ago in Kosti city in the middle of
Sudan, by the White Nile River (Fig. 1). The building was
designed and constructed over an area of 1350 sq. m toprovide advanced medical services for the citizens of theWhite Nile state. Because of inadequate geotechnicalinvestigation prior to construction, the building wasconstructed in an area of highly potential expansive soils,
without any special considerations for this type of soil.
The building is a R/C frame type building, the frame is acolumn-flat slab type, supported on isolated footings.
All parts of the building are used as rooms for doctors
and hospital wards for patients.
As reported by the local governmental authorities, a few
years after completion of construction of the hospital, thebuilding started to suffer from cracks and visible movementsof the expansion joints. Lack of routine inspection andmaintenance and poor previous rehabilitation works had led
to continuous and fast development of cracks andmovements.

Expansive soils are soils that have potential for swelling andshrinkage under changing moisture conditions. The volume
change resulting from swelling and shrinking causes damages
to structures founded on them. The expansive soil area
includes nearly all the agricultural schemes and most of thedevelopment projects in Sudan and covers about 40% of thetotal area of Sudan (Osman and Charlie 1983).
Damages of structures caused by expansive soils have beenreported from different locations in the clay plain (Lates et al1983). The damages include buildings, roads, factories,
hydraulic structures etc., and were attributed to lack of proper
identification and classification of expansive soils andimproper design of the foundations of the damaged structures(Elsharief, 2009).
3. ASSESSMENT PROGRAM AND RESULTS
A comprehensive structural-geotechnical assessment
program was designed and implemented. The following
sections discuss the steps and findings of this program.
3.1 Visual inspection & excavations
A careful visual inspection was performed for the structuralmembers of the building, it was clear from the inspection thatthe building was unstable since expansion joints were
undergone excessive rotations (Fig. 2) and severe structural
cracks were seen in some columns. The maximum openingof the expansion joints at the top of the building was about
13cm which indicates that the building had undergone a totalrotation of 0.67 degrees. Fortunately, foundations were notaffected by these movements. A number of internal andexternal footings were excavated for visual inspection andmeasurements. Excavation has been done on some of
the foundations with great difficulty because of the saturation

Figure 1. Location of Kosti Military Hospital (Google Earth).
70

of the clay soil, and in some cases, water
gushed profusely about the footings (Fig.3). Because of theupheaval movement of the soils, and probably due to theweakness of concretes some of the basement columns were
buckled and crushed (Fig. 4). All sanitation systems of thebuilding were damaged. Leakage and accumulation of waste
water and chemicals (waste of hospital’s laboratories) aroundthe columns and in contact with the basement cover slab had
caused deterioration and corrosion of reinforcement ofcolumns and slabs (Fig. 5).
Figure 5. Corrosion of reinforcement of basement slab andcolumns
3.2 Detailed geotechnical investigation
3.2.1 Drilling of Boreholes
Three boreholes were drilled, two of them to about 10.0m andthe third one to 15.0m depth, in different locations around thebuilding. The boreholes were drilled using a rotary soil
mechanics drilling rig. Disturbed soil samples were collected
at 1-m meter intervals for visual inspection and classification
tests. Undisturbed samples were taken at 1.5-m intervals fromthe cohesive soil. The standard Penetration Test (SPT) was
performed in silt or sand layers or whenever there wasdifficulty in getting undisturbed soil samples. The SPT was
performed by driving an open tube sampler (50mm O.D.) by
the blows of a standard automatic hammer weighing 64kg andfalling freely from a height of 760mm. The number of blows
required to give tube penetration of 300mm was taken as
SPT-N value.

Figure 2. Opening at expansion joint.

3.2.2 Excavation of Test Pits
Three test pits were excavated manually to about 4.0m at
different locations around the building. The disturbed soil
samples were collected at each meter interval for visualinspection and laboratory testing.

Figure 3. Dewatering during excavation.

3.2.3 Groundwater
No groundwater was observed during the drilling of
boreholes in this site. It is expected to be far below theeffective zone of the foundations.
3.3.4 Soil Conditions
The soil profile for the two boreholes and test pits is shown inAppendix (A). The three boreholes showed relatively similarsoil stratification consisted of dark brown silty clay of highplasticity (CH) and dark brown to yellowish greyish veryloose to medium dense clayey sand (SC) at the top 4.0m to
5.5m. Below this alternative layers of very stiff brown low
plastic silty clay (CL) and very stiff brown low plastic silt(ML) and very stiff dark to light brown highly plastic silty
clay (CH) were encountered down to the bottom of boreholes
at 10.0m and 15.0m depths.

Figure 4. One of buckled basement columns

71Test pits soil profile showed one layer of dark to light brownsilty clay of high plasticity (CH) extending down to thebottom of test pits at 4.0m depth.

3.2.5.7 Swelling Pressure Test
The swelling pressure tests were conducted on undisturbed
soil samples. The volume of the soil was kept constant in an
oedometer cell, while the soil was saturated with water during
the test. The results are shown in Table 1 below. The results
reflected low swelling pressures. This is due to high moisture
content of these samples, usually high swelling pressuresdepend on high moisture content deficiency. Also the soil
samples may be subjected to some disturbance during
mobilization.

3.2.5 Laboratory Testing
Laboratory tests were performed on a number of disturbed
and undisturbed soil samples. The tests included Atterberg
limits, grain size analysis, natural moisture content, UU
triaxial test, consolidation test, swelling pressure test andchemical tests. The procedures followed were in general
conformance with those recommended in the British Standard
BS 1377 (1990). Soils were classified according to the
Unified Soil Classification System (USCS).

Table 1: Swelling Pressure Test Results
InitialMoisture
Content (%)

Final Moisture
Content (%)

Soil
Type
According to
USCS

3.2.5.2 Atterberg Limits
The tests were carried out on clayey soil samples taken fromdifferent depths. The results gave liquid limit values ranging
between 29% to 85% and plasticity index values between11%
to 60%. This indicated high swell potential for the clayey soilat the present site.

1.5
3.0
3.0
3.0

Swelling
Pressure(kN/m²)1
123

Depth (m)B.H. No.

3.2.5.1 Classification Tests
The objective of these tests was to reveal soil types
encountered at different depths of the boreholes. Tests carriedout included Atterberg limits, grain size analysis and natural
moisture content. These classification tests confirmed the soil
profile described in subsection 3.5.

0
20
20
20

29.05
19.05
21.92
23.70

33.68
23.23
24.80
24.52

CH
SC
SC
CH

3.2.5.8 Chemical Tests
Chemical tests were performed according to BS 1377,1990-2
on two soil samples and one water sample to determine
chloride content, sulphate content and the pH value. The
results are shown in Table 2 below. The results indicatedalkaline soil and water of low contents of chloride andsulphate. Therefore, the building foundations are notexpected to be affected by these low contents of chemicalsalts.

3.2.5.3 Grain Size Distribution
The grain size distribution of the soil samples was determinedin the laboratory. The results of this test with combination of
Atterberg limits test results were used to assess the soil types
according to the Unified Soil Classification System (USCS).

Table 2: Chemical Test Results
Location

Depth (m)Chloride(%)

Sulphate(%)

3.2.5.6 Consolidation Tests
Consolidation tests provide information for use in evaluating
the compressibility of the soils and estimating the settlementof foundations established on these soils. The consolidation
tests were performed on undisturbed clayey samples. First,
the swelling pressure test was conducted on the samples.
Then the samples were loaded beyond the swelling pressure
to allow consolidation to proceed.

SampleType

3.2.5.5 UU-Triaxial Test
The undrained unconsolidated (UU) triaxial test was
conducted to measure the shear strength parameters of thesoil. The average measured shear parameters (Ø & C) were
9.33º and 45.93 kPa respectively.

Sample No.

3.2.5.4 Natural Moisture Content
The natural moisture content was determined fromundisturbed samples and by weighting the disturbed samples
obtained from SPT sampler in the field and then the test wascontinued in the laboratory.

pH

1

Soil

B.H. 22.0

0.21420.2513

5.27

2

Soil

B.H. 32.0

0.22130.1642

4.34

3

Water

-

-

0.20570.20435.10

3.2.6 Analysis of geotechnical results
It is evident from the soil profile that the top 4.0 to 5.5m layerof highly plastic silty clay (CH) and very loose to medium
dense clayey sand (SC) was encountered. This is underlainedby alternative layers of very stiff low plastic silty clay (CL)and very stiff low plastic silt (ML) and very stiff highly
plastic silty clay (CH) extending down to the end of
boreholes. No groundwater was encountered in boreholes
during this investigation; it is expected to be far below theeffective zone of the foundations. Although low values of72swelling pressures were measured in the laboratory, the index
properties of the investigated soil indicated high swelling
potential for the clayey layers. The SPT N-values generallyindicated soft to very stiff clays and very loose to medium
dense sands, which reflects low to medium shear strength forthe soil at the present site. The chemical test results showed
acidic soils of high contents of harmful chemical salts. The
building was constructed on isolated footing foundations
placed at about 2.9-m depth below the ground surface level.
At this depth the bearing capacity of the soil recommended to
be 140 kN/m2.
The bearing capacity determination was based on soil
classification and SPT results. Use SPT-N average value
below the foundation level and maximum footing width of
4.5m for permissible settlement of 25mm at (Terzaghi andPeck Chart, Bowles, 1994).

practicable in testing sound concrete. Especially ininvestigation of crack depth, it is ineffective if the crack is
water filled. The performance is also often poor in very roughsurfaces. Sometimes good contact requires the use of acoupling gel between the transducers and the structure. Soundenergy above the audible frequency of 16,000 Hz is
designated as ultrasonic (Song & Sarthawathy, 2007). The
instrument consists of transmitter and receiver (two probes).
The time of travel for the wave to pass from the transmitter tothe receiver when kept opposite to each other is recorded in
the ultrasonic instrument. The distance between the twoprobes (path length) can be physically measured. Henceultrasonic pulse velocity = path length/time. This velocity inconcrete can be related to its compressive strength. Different
approaches have been published for classification of concrete
quality based on the pulse velocity. In this study theWhitehurst (5) classification method was adopted. According
to the Whitehurst classification method, the quality ofconcretes of the columns of different floors was found to bevarying between poor (pulse velocity < 3Km/sec.) anddoubtful (pulse velocity < 3.5 km/sec.), whereas some points
were found to indicate very poor quality (pulse velocity <2.5km/sec.).

3.3 Field and Lab. Testing of concretes
A combination of destructive core tests, ultrasonic pulse
velocity and rebound hammer tests was performed to find thecompressive strength and to assess the quality and uniformityof the concretes of the building.
3.3.1 Core test results
Six core specimens were extracted from the foundations andfloor slabs using an electrical type core cutting machine. All
core specimens were thoroughly inspected and prepared forthe compression test. Visual inspection of the core specimens
before compression tests revealed the following:- Entrapped air voids were noticed in the core specimens
which indicated that the concretes of different structuralelements of the building were not properly vibrated andcompacted.
- The concrete mix was found to be over sanded (i.e. theamount of cement was less than required).
- Many decayed aggregate particles were were noticed in
the mix. Also the maximum size of aggregate found wasless than 15mm and the grading observed was not
uniform.
- After visual inspection the specimens were prepared andtested under uniaxial compression load, the averagevalue of the strength results was 14N/mm2.

3.3.3 Rebound hammer test results
This test is basically a surface hardness test and should be
used only where the surface has not carbonated as the results
tend to be very high and unrealistic on carbonated surfaces. It
is also possible to evaluate the degree of uniformity ofconcrete through statistical analysis of test results (ACI,1195).
The building under consideration was constructed four years
ago so little carbonation effect is possible. Rebound hammer
values obtained from different columns at different floor
levels showed low surface hardness (average rebound number14) and high variability (standard deviation of reboundnumber 12.5) in the properties of concretes of columns.
3.4 Structural analysis
A structural analysis of the concrete frame was performed
based on the actual dimensions of the building and main
findings of field and laboratory tests results. The structural
analysis showed that all the basement columns, second floor
columns and the basement cover slab were stressed beyond
the ultimate limits. While the bearing capacity of foundations
were found to far less than the applied axial and bending
forces of basement columns, the uplift pressures of soil didn’t
affect the punching shear strength of foundations.

3.3.2 Ultrasonic pulse velocity tests results
Ultrasound pulse velocity was performed to assess the qualityof concretes of columns and to evaluate the degree ofcompaction.
Ultrasonic pulse velocity test is usually performed forassessing the quality and integrity of concrete by passing
ultrasound waves through the specimen or reinforced
concrete member under test, This test can also be used todetermine the presence of honeycombs, voids, cracks etc. The
main strength of the method is in finding general changes in
condition such as areas of weak concrete in a generally soundconcrete. At the same time, the UPV technique is not always

4. SUGGESTED STRENGTHENING METHODS
After a comparative socio-economic feasibility study of thetwo options of demolishing and strengthening, it was decided
by the different consultancies and authorities involved tostrengthen the building. Time was critical since the hospitalbuilding had to be restored to provide medical services. Based73reinforcements were fixed around the columns before final
concreting.

on the findings of the assessment program, it was clear thatthe building required faster major and minor strengthening
and retrofitting. It was also critical for building andstrengthening works to be implemented before the rainyseason, so as to prevent any moisture variation in soils and tostop any further movement of the building skeleton. Five
different major structural strengthening techniques weresuggested and implemented.

4.3 Metal jacketing of columns
In order to increase the load carrying capacity of the second
floor columns while maintaining the same cross-section,
metal jacketing was the best solution. The columns were
strengthened using four channel sections at the four corners
along the columns, connected via horizontal steel strips at asuitable spacing (Fig. 7). The gaps between the concrete andthe added steel frame were filled with a suitable epoxy
mortar.

4.1 Modification of foundation system
It was suggested that the isolated footings of the foundation to
be strengthened and connected using reinforced concretebeams fixed on top of footings. This will increase thestiffness and tightness of the foundation; hence redistributethe bearing stresses and the non-uniform uplift soil pressures.
Also this will eliminate any stress concentrations and prevent
any further increase in differential movement of underneathsoils (Fig.6). This was performed after restoring the affectedfootings to their initial levels. Soils beneath the footings
affected by the uplift pressures were removed and thefootings were allowed to restore gradually to the initial levels
under the existing dead loads. Wherever needed, additional
surcharge loads were applied to push the footings down totheir initial levels. For damaged short columns, shoring ofupper floors was sometimes needed to reduce the applied
loads, strengthen the columns and restore the footings.

Figure 7. Metal jacketing of second floor columns
4.4 CFRP for basement cover slabs
The basement cover slab (total area 1350 sq.m) was
strengthened using carbon fiber reinforced polymer laminates
fixed at the soffit of the concrete slab using selected epoxy
resin (Bank, 2006). This technique was the best solution to
remedy the two deficiencies of the slab; structural cracks due
to foundation movements and corrosion of reinforcements
due to chemical attack (Fig. 8).

Figure 6. Modification of foundation and concrete jacketingof short columns
4.2 Concrete jacketing of short columns
Strengthening of basement columns using concrete jacketing
was apparent since the load carrying capacities of thesecolumns were much less than the applied axial forces andbending moments. Shear connectors were first inserted andfixed in the old concrete columns using a carefully selected
epoxy binder. The surfaces of the old concrete columns wereroughened and a suitable epoxy binder was distributed before
the additional reinforcement was placed (Hollaway andLeeming, 2005). After that additional flexure and shear

Figure 8. CFRP for basement cover slabs

74

4.5 Shear wall installation

6. REFERENCES

To enhance the stability of the building and to rectify the
openings at the expansion joints caused by the differential
movement of the foundations, concrete shear walls were
installed at both sides of each expansion joint along the totalheight of the building (Fig. 9).

[1] Mohammed A. Osman & W. A. Charlie, (1984).
“Engineering Properties of Expansive Soils in
Sudan” 5th International Conference on Expansive
Soils, Adelaide, South Australia, 21-23 May 1984,
pages 311-315.
[2] Lates E. M., Elmonshid B. E. F. and Abbo S. H.,(1983) “Review of Problems Generated by
Expansive Soils in Sudan” Proc. of Seminar on
Expansive Clay Soils Problems in Sudan, Wad
Medani, January 1983, Hydraulic Research Station,
HRS.
[3] El-sharief A. M., (2009). “Foundations on
Expansive Soils Sudan Experience” Proc. Of First
Conference of Postgraduate Studies and Scientific
Research, University of Khartoum, Sudan.
[4] Joseph E. Bowles, (1994) Foundation Analysis andDesign, Fifth Edition.
[5] Ha-Won Song & Velu Sarthawathy (2007),“Corrosion Monitoring of Concrete Structures- A
Review”, Int. Journal of Elect. Chemical Science,2(2007) 1-28.
[6] American Concrete Institute, (1995), “In Place
Methods to Estimate Concrete Strength”, Reported
by ACI Committee 228.
[7] L.C.
Hollaway&M.
B.
Leeming(2005),”Strengthening of Reinforced Concrete
Structures”, CRC Press, Woodhead Publishing
Limited, England.
[8] Lawrence C. Bank (2006).”Composites for Con
struction-Structural Design with FRP Materials”,
John Wiley & Sons, INC.

Figure 9. Installation of shear walls at expansion joints
5. CONCLUDING REMARKSThis case study highlights the following facts:- Clever combination between different strengthening andretrofitting techniques can save a lot of money and time.
- When considering the strengthening and rehabilitation of
structures, it is essential to establish the nature of the
problems, extent, severity and exact causes. If this is notdone the symptoms could be treated rather than thecauses.
- There is no substitute for following sound design andconstruction principles when constructing structures on
expansive soils.
- Careful and continuous monitoring during and after
installation of strengthening and retrofitting works isvital to evaluate the efficiency of such works.

.

75

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Gravitational Sphere Packing Method to Find the Porosity of the Soil Particles
with the Identical Size inside the Cylindrical Container
Mohammad Mahdi Roozbahani , Bujang B. K. Huat, Afshin Asadi
Universiti Putra Malaysia
mahdi_roozbeh@yahoo.com, bujang@eng.upm.edu.my, afshin.asadi@yahoo.com,
Civil Engineering Department, Faculty of Engineering, Universiti Putra Malaysia, 43400 UPM-Serdang,
Malaysia
ABSTRACTThe present article demonstrates new gravitational sphere packing with geometrical method within the cylindricalcontainer. Wall effects are considered for this container to simulate the reality conditions in the laboratory works. It isshown that the porosity is reduced by increasing in the volume size of the container, and the same results of the porosity
have been achieved for both mathematical and geometrical methods.
Keywords: Porosity, wall effects, sphere packing, cylindrical container, gravitational sphere packing

have expressed periodic boundary condition to eliminate
the wall effects and provide the raw data of porosity [12].
Periodic boundary condition is a computational box whichis repeated infinitely in x and y direction .In fact, periodic
boundary condition checks on the spheres which are exitedfrom the wall, and those spheres enter inside the containerfrom opposite side.
In some cases, the containers are not eliminated tosimulate the reality condition of the interaction of theparticles with each other, thus the resulted porosity would
have higher value in comparison with the manner without
the effect of wall. The higher porosity results will beachieved without wall effects because of the partial
ordering of soil particles near the wall. Consequently, themost porous regions of a container are near the walls, but itshould be mentioned that shape of the containers have someeffects in partial ordering.
The most common containers are cylindrical and squarecontainer. Of these, cylindrical container is more popular
because of the having same shape with the equipments inlaboratory.
Mathematical [13] and geometrical [10] are the mostcommon ways to simulate the sphere packing. There arepros and cons for the both way. In the former one, theporosity can be calculated by mathematical formulas
analytically, whereas it is not able to depict the spheres andinteraction of them. In the latter one, the speed of packing
process will decrease by increasing the number of elements.
The macro and micro structures of particles, however, canbe defined minutely. The geometrical approaches areapplied for this present paper, therefore different
geometrical methods are employed to create the structure ofloose sphere packing. Apollonius 3D has been
implemented for equations of tangents controlling
treatments of soil particles.

1. INTRODUCTION
Sphere packing is a method which has been used inmany fields such as geotechnical and chemical engineering,
physic, ceramics, powder technology, and biology [1][2]-[3]-[4]-[5]-[6]-[7]-[8]-[9]. Porosity calculation of thesoil is important from the aspect view of finding the void
fraction of the particles, permeability computation, andinteraction between the particles.
There are two types of sphere packing which are 1:
loose or gravitational sphere packing, and 2: close spherepacking. The former one simulates the conditions of theunconsolidated soil particles with loosely interior structure,
whereas the latter one simulates the conditions ofconsolidated soil particles with fully compacted particulatematerials. The model structures of two sphere packing arequite different with each other. In the first one, the methodof the drop and roll from specified height is applied, while
for the second one, providing the most compacted packed
bed is the striking aim which should be met.
In the current paper, authors utilized the loose or
gravitational sphere packing which can simulate layering of
the soil particles without any compaction called
unconsolidated porosity value. Tory was the inventor ofthis method, whilst the other researchers have promoted thework of him [10]-[11]-[12]. Loose sphere packing can
employ equal-sized spheres from nano- till coarse-sizedparticles, hence wide variety applications of this method
would be very helpful in the field of soil particles
interaction.
Variations of container size have positive correlation
with the final porosity results. However, some researchers762. Computational method

Stoyan found the porosity values of loosely packed
spheres by the means of “greedy algorithm”.
The porosity results are decreased by increasing thediameter and the height of the container, and the reason isinterpreted by the effects of the container volume on the
results.
One sample of sphere packing is shown in Fig. 1, 2,
and 3 which are the 3D and 2D view of packed spheres.
Container shape is not completely shown in order to better
observation of the identical size of the soil particles, andinteraction with them.

The Tory’s method has been applied for the current
work. However, pure geometrical approaches have been
employed rather than vectors manipulations. The base of
the work of Tory can be summarized as follows:
1: The initial co-ordinates of spheres are chosen based on
the random number generator on which the Monte-Carlo
method is implemented.
2: The first height of every sphere is chosen based on thepreviously settled particles, hence the upmost height ofpre-settled spheres is chosen, and the first height ofincoming sphere is defined according to the upmost height.
3: The incoming sphere may collide with other fixed
spheres. In this case, the stability conditions have to becontrolled for the incoming sphere. If there have been threespheres underneath of the incoming sphere, the sphere isdetermined as a stable sphere; otherwise that sphere rolls
down on the other spheres to find the stability condition
which is usually called reaching the minimum potential
point. Subsequently, there are two stability methods. The
first one is being stable on the top of the other spheres andthe second one is being stable by hitting the floor [10]-[11]-[14].
Apollonius 3D governs the equations of the contacting
an incoming sphere with one, two, and three spheres. The
formulas for contacting with three spheres are as follows:
Take for example:
The fixed spheres have been determined with 1, 2 and 3
subscripts and incoming sphere is determined by “s”
subscript.

Figure 1. Three dimensional view of the sphere packingmodel(𝑋𝑠 − 𝑋1 )2 + (𝑌𝑠 − 𝑌1 )2 + (𝑍𝑠 − 𝑍1 )2 = (𝑅𝑠 + 𝑅1 )2 (1)(𝑋𝑠 − 𝑋2 )2 + (𝑌𝑠 − 𝑌2 )2 + (𝑍𝑠 − 𝑍2 )2 = (𝑅𝑠 + 𝑅2 )2 (2)(𝑋𝑠 − 𝑋3 )2 + (𝑌𝑠 − 𝑌3 )2 + (𝑍𝑠 − 𝑍3 )2 = (𝑅𝑠 + 𝑅3 )2 (3)3.

Results and discussion
The simulated sphere packing is verified with theliterature, and the verification has been done with themathematical method [15], which can explore that theresults of the mathematical and geometrical are same
intrinsically. The results are shown in Table 1.
Table 1 Comparison of porosity between geometrical andmathematical method.
Height ofContainerthe
Porosity2 Porosity3
1N
size(radius,
container,
(%)
(%)
mm)
mm)
50
3.5
10.6177
54.7885
54.15
60
3.5
12.6174
53.393653.7
80
4.5
9.662052.9012
52.63
100
4.5
11.9555
51.5812
51.31
125
4.5
14.7689
50.992050.15
1: N is the number of the spheres as to the literature.
2: Porosity results of present work.
3: Porosity results in literature.

Figure 2. Cross section sample of the sphere packing

77

Figure 3. Plane view of the sphere packing

4.

ACKNOWLEDGMENT

Support for this work by the University of Putra
Malaysia Research Board is grate-fully acknowledged.
5.

CONCLUSION

It is shown that the values of porosities for the bothgeometrical and mathematical methods are very close toeach other.
It is construed that with increasing in the volume of thecontainer, the porosity will be decreased subsequently.
Because, there are more porous regions near the wall of thecontainer, and increasing the volume size of the container
will diminish these effects drastically.
6. REFERENCES[1]

K. Sobolev, A. Amirjanov, Application of genetic algorithm formodeling of dense packing of concrete aggregates, Construction
and Building Materials. 24 (2010) 1449-1455.

[2]

A.V. Kyrylyuk, A. Wouterse, A.P. Philipse, Random Packings
Of Rod-Sphere Mixtures Simulated By Mechanical Contraction,
Physics. (2009) 211-214.

[3]

G. a Georgalli, M. a Reuter, A particle packing algorithm forpellet design with a predetermined size distribution, Powder
Technology. 173 (2007) 189-199.

[4]

L.T. To, Z.H. Stachurski, Random close packing of spheres in around cell, Journal of Non-Crystalline Solids. 333 (2004)
161-171.

[5]

C.R. a Abreu, R. Macias-Salinas, F.W. Tavares, M. Castier, A
Monte Carlo simulation of the packing and segregation of
spheres in cylinders, Brazilian Journal of Chemical Engineering.
16 (1999) 1-12.

[6]

a Galakhov, V. Shevchenko, Analysis of two-dimensional
packings, Ceramics International. 18 (1992) 213-221.

[7]

A.N.E.W. Algorithm, F.O.R. Close, R. Packing, Computer
simulation of close random packing of equal spheres’, PhysicalReview A. 32 (1985) 2347-2351.

[8]

T. Quickenden, Random packing in two dimensions and thestructure of monolayers, Journal of Colloid and Interface
Science. 48 (1974) 382-393.

78

[9]

D. Haughey, Local voidage variation in a randomly packed bedof equal-sized spheres, Chemical Engineering Science. 21(1966) 905-915.

[10]E.M. Tory, B.H. Church, M.K. Tam, M. Ratner, Simulated
random packing of equal spheres, The Canadian Journal ofChemical Engineering. 51 (1973) 484-493.

[11]

W.S. Jodrey, E.M. Tory, Simulation of random packing of
spheres, Simulation. 32 (1979) 1-12.

[12]

Y. Shi, Y. Zhang, Simulation of random packing of sphericalparticles with different size distributions, Applied Physics A. 92(2008) 621-626.

[13]

Y.G. Stoyan, G.N. Yaskov, Packing identical spheres into acylinder, International Transactions in Operational Research. 17(2009) 51–70.

[14]

W.S. Jodrey, Computer Simulation of Equal Spheres ofIsotropic , Homogeneous , Dense Random Packing, 30 (1981)
111 - 118.

[15]

Stoyan, Yu.G, Mathematical methods for geometric design,
Proceedings of International Conference PROLAMAT 82.
(1983) 67–86.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 19-21, 2012, ISBN: 978-4-9905958-1-4 C3051

Parametric Study of Piled Raft Foundations to Improve the Design Method
Abbasali Taghavi, Asskar Janalizade Choobbasti
Babol Noshirvani Institute of Technology, Iran
ABSTRACTThe piled raft Foundation is a geotechnical composite construction, consisting of pile, raft and soil elements, which isapplied for the foundation of tall buildings in an increasing number. Conventionally, there are two main design
methodologies for this kind of foundation. The first design method is based on reducing raft foundation settlement byadding a certain number of piles. Another design method assumes that all structure loads to be transferred to piles
without considering contribution of load taken by the raft. In this paper a parametric study of piled raft foundations,taking account of their components interaction, accomplished by Finite Element Method (FEM). The results show thatconventional methods of design are based upon relatively conservative assumptions and using of numerical methods
considering interaction between the piled raft components, might give an economic design approach.
Keywords: Piled raft foundation; Settlement – based design; Finite element method; Soil – Structure Interaction
Allowances were made for the piles to reach their ultimate
capacities and local bearing failure of the raft. Russo [10]employed a similar method where the piles and soil weremodelled by linear or non-linear interacting springs. The soil
displacements were calculated using Boussinesq’s solution
thus yielding a closed form solution. The non-linear
behaviour of the piles was modelled by the assumption of a
hyperbolic load-settlement curve for a single pile. Brown and
Wiesner [11] used the Boundary Element Method to analyse astrip footing supported by equally spaced identical piles
embedded in an isotropic homogenous elastic half space.
Application of Mindlin’s solution was used to determine theinteraction relationships due to the interface forces. Poulos[12] extended a method of BEM to incorporate the effect of
free-field soil movement, load cutoffs for the pile - soil andraft – soil interfaces to examine the interaction mechanism
between the piled raft and a soil subjected to externallyimposed vertical movement. The analysis is implemented via
a computer program PRAWN. Hooper [5] used Finite
Element Method to analyze a piled raft and considered thestiffening effect of the superstructure on the behaviour of the
piled raft by introducing an equivalent raft thickness. Liu andNovak [13] employed the finite element method to examine
the behaviour of a raft supported by a single pile at the centre.
Mapped infinite elements are used to model the far - field soil
medium to simulate an unbounded domain. A weak zone
around the pile with lower strength and modulus wasintroduced to account for the slip at the pile-soil interface.
Reul and Randolph [14] presented a three-dimensionalelasto-plastic finite element method using program ABAQUS
for the analysis of piled raft foundations in overconsolidated
clay. Oh etal [15] used finite element program PLAXIS for
the analysis of piled rafts on Australian sand.
Combined Boundary and Finite Element Method as well as
Combined Finite Element and Finite Layer Method are used
to reduce the calculation time by some authers including
Mandolin and Viggiani [16] and Zhang. In this paper finite
element is employed to perform parametric studies.

1. INTRODUCTION
In the past decade, piled raft foundations have been widelyused in practice to support heavy and high - rise buildings andextensive studies have been done about their design methods.
The piled raft foundation is a combination of a raft foundation
with piles. Taking into account the design requirements andbearing capacity, piles in piled rafts are designed. In soft soils
where a raft alone may not provide the bearing capacity,
connected piles to enhance the performance of the raft or
unconnected piles to reinforce the supporting soil are usually
added. The most effective application of piled rafts occurs
when the raft can provide adequate load capacity, but thesettlement or differential settlement of the raft alone exceeds
the allowable values. In this situation, piles are used as
settlement reducer and designed to have small size andnumber in comparison with conventional pile groups.
Piled raft foundation consists of three carrying elements: soil,
raft and pile, which the interaction between the elements may
cause some complexity in prediction of this composite
foundation.
One of the Technical Committees of the International Society
for Soil Mechanics and Foundation Engineering (ISSMFE)
focused its efforts in the period 1994 - 7 towards piled raftfoundations, collected considerable information on casehistories and methods of analysis and design, and produced
comprehensive reports on these activities including O'Neilletal. [1]; van Impe & Lungu [2]. The concept of piled raftfoundations is by no means new, and has been described byseveral authors, including Zeevaert [3], Davis & Poulos [4],Hooper [5], Burland et al. [6], Sommer et al. [7] and Franke[8], among many others.
Poulos [9] developed a program GARP which employs anapproximation method for the analysis of piled raft with theconsideration interaction effects between the piles and raft.

79

compute the detailed distributions of settlement, bending
moment and shear in the raft, and the pile loads and moments.
The first and second stages involve relatively simple
calculations, which can usually be performed without acomplex computer program. The detailed stage will generally
demand the use of a suitable computer program that accounts
in a rational manner for the interaction among the soil, raftand piles and may be the effect of the superstructure. In orderto do this stage carefully, there are many approaches
including “strip on springs” approach, “plate on springs”
approach, boundary and element method and etc. In this paper
we employ mentioned design process and philosophies toperform a suitable design.

2. DESIGN CONCEPTS
As with any foundation system, the design of a piled raftfoundation requires the consideration of a number of issues,
including: ultimate load capacity for vertical, lateral andmoment loadings, maximum settlement, differential
settlement, raft moments and shears for the structural designof the raft, pile loads and moments, for the structural design ofthe piles.
In much of the available literature, emphasis has been placedon the bearing capacity and settlement under vertical loads.
While this is a critical aspect, and is considered in detailherein, the other issues must also be addressed. In some cases,
the pile requirements may be governed by the overturning
moments applied by wind loading, rather than the verticaldead and live loads.

3. FINITE ELEMENT MODELING
3.1 Finite Element Mesh and Boundary Condition
The behavior of the piled raft was investigated by carryingout 3D numerical analyses. The geotechnical finite element
code PLAXIS (Vermeer and Brinkgreve [19]) was used forthis purpose and 15-node triangular elements were used. Fig.1
shows a typical finite element mesh used for parametricstudy. This parametric study has been done for different raftand pile sizes and different number of piles. Material
properties of soil and piled raft are shown in Table I.

2.1 Alternative Design Philosophies
Randolph [17] has defined clearly three different design
philosophies with respect to piled rafts:(a) The “conventional approach”, in which the piles aredesigned as a group to carry the major part of the load, while
making some allowance for the contribution of the raft,
primarily to ultimate load capacity(b) “Creep piling”, in which the piles are designed to operate
at a working load at which significant creep starts to occur,
typically 70 - 80% of the ultimate load capacity; sufficient
piles are included to reduce the net contact pressure between
the raft and the soil to below the preconsolidation pressure of
the soil.
(c) Differential settlement control, in which the piles arelocated strategically in order to reduce the differential
settlements, rather than to reduce the overall average
settlement substantially.
2.2 The Design Process
Poulos [18] has suggested that a rational design process for
piled rafts involves three main stages:(a) A preliminary stage to assess the feasibility of using apiled raft, and the required number of piles to satisfy design
requirements
In the preliminary stage, it is necessary first to assess theperformance of a raft foundation without piles. Estimates ofvertical and lateral bearing capacity, settlement anddifferential settlement may be made via conventional
techniques.
(b) A second stage to assess where piles are required and thegeneral characteristics of the piles
Much of the existing literature does not consider the detailed
pattern of loading applied to the foundation, but assumes
uniformly distributed loading over the raft area. While thismay be adequate for the preliminary stage described above, it
is not adequate for considering in more detail where the piles
should be located when column loadings are present.
Assessment of the maximum column loadings for this control
is the subject of second stage.
(c) A final detailed design stage to obtain the optimum
number, location and configuration of the piles, and to

Fig1. Typical Finite Element mesh used for numericalmethodTable I. Material properties used for analysisnE (MPa)γ (kN/m3) C (kPa)
Soil
14
0.18
18
5
Pile
25000
0.20
24
Raft
25000
0.20
24
-

φ

30
-

The pile head was connected to the raft rigidly. Special
interface element was used at the soil – pile and raft – pileinterface. A relatively fine mesh was used near the interfaces
while a coarser mesh was used further from the pile and raft.
Vertical boundaries are assumed to be on rollers to allow
downward movement of the soil layers and the beneath of themodel is fixed in three orthogonal directions. For the far-fieldboundaries, the distance of the boundary from the edge of theraft was greater than the pile length since the observed
influence zone based on the finite element analysis includinginterface is there [20]. This limit prevents the effects of80

boundaries and works as infinite elements to simulate half
space. After initial equilibrium, the vertical loading wasapplied on the top of the raft surface. Since modeling of theentire pile installation process is rather complicated, the pile
was assumed to be in a stress-free state at the start of theanalysis.

Table II. Comparison of the results with the test data
Average Settlement, Load taken by
Savg (mm)piles (%)

3.2 Constitutive Modeling
Based on the materials, it is preferable to use theMohr-Coulomb model for relatively quick and simple andfirst analysis of any problem considered. To simplify the
analysis process, average values of material parameters (asmentioned in Table I) were adopted for the soil layer. The raftand piles has great Young’s modulus in comparison with thesoil and remain in elastic area, so they were modeled with alinear elastic model.
The modeling techniques used for the pile - soil interface are
generally divided into two types: slip element and thin layer
element. The former was used by Jeong etal. [20] and Lee
etal. [21], in which the slip behavior could be considered. The
latter was used by Reul and Randolph [14] and de Sanctis andMandolini [22]. In this paper thin layer elements with athickness of 0.1d (d, pile diameter) is used.

5. PARAMETRIC STUDY

Present Study

21.67

23.55

P

Fig.3 shows different arrangements used for parametricstudy. Arrangements have 16, 25, 36, 81 and 169 piles,
respectively. Arrangement 3 and 5 have s/d= 4 and others
have s/d=6.

Arrange 1

Arrange 2

Arrange 3(2S center + S corner )(1)
where S corner is the settlement of the raft corner and S center isthe displacement of the raft center of piled raft system.
3

19

Table III. System configuration for parametric study
Length40 m – combined 20 and 40 m
0.5 m
DiameterPiles/d= 4 , 6Distance
Center- Uniform
Arrangement
0.5, 1, 2 m
ThicknessRaft
26 by 26
Width
150 , 200 kN/m2
Uniform
Load
Equal to uniform load (on piles)
Concentrated

The validation of the present model was done by acomparison with a centrifuge model test for a circular piled
raft on stiff clay, which was carried out by Horikoshi and
Randolph [23]. The total number of piles was nine (a 3 × 3
array with a pile spacing of 2.5 m) with a length of 15 m anddiameter of 0.32 m, which were located under a 14 m
diameter circular raft with a thickness of 0.05 m. A soil depthof 25 m was taken into account. Eq.1 describes the average
settlement. Also a comparison of the results with a numericalmethod proposed by Prakoso and Kulhawy [24] has been
done. The results are shown in Fig.2 and Table II.
122

Table III lists soil, raft and pile configurations, different sizes
and distances and number of piles used in the raft only andpiled raft analyses.

4. VALIDATION

S average ≈Horikoshi and
Randolph

Arrange 4
Arrange 5
Fig3. Different arrangement used for parametric study
In order to consider more detailed design stage, as mentioned,
uniform loading results (preliminary stage) are comparedwith the nonuniform and concentrated loading one. For aconcentrated load, the load was applied at the center of theraft, but the equivalent magnitude of the load in the twodifferent loads was the same. The vertical loading wasapplied on the top of the raft surface after initial equilibrium.
All analyses were carried out under drained conditions. Thus,
hydrostatic water pressure distribution was assumed in adrained condition with the ground water table located on the
top of the soil layer.
6. COMPUTED RESULTS AND DISSCUSION
As mentioned before, maximum settlement, differential
settlement, raft moments and pile loads are the main

Fig2. Comparison between present method and a numericalmethod81

parameters in the assessment of the piled raft behaviour. Also
concentrated and local loading is used to consider the design
stages proposed by poulos in addition to uniform loading.
Fig.4 shows the variation of the normalized average
settlement with the square root of the number of piles, n p .

normalized average settlement with the square root of thenumber of piles, n p . The normalized differential
settlement, S d , is calculated from Eq.2.

∆wc −e E s(2)q ⋅ Brwhere ∆w c-e is the differential settlement, E s is the modulus ofelasticity for the soil, q is the applied load and B r is raft
breadth.
In this case, arrangement 3 does not act so well and shows
significant differential settlement especially underconcentrated loading. As expected, when the load applied incertain point (on the piles), significant local displacement
occurs and make remarkable differential settlement.
Sd =Each value of this parameter also introduces an arrangement.
The average settlement is calculated using Eq.1 andnormalized respect to average settlement of the raft only.

R

R

R

R

R

R

Fig4. Variation of normalized average settlement with thesquare root of the number of piles (a) Load: 150 kPa (b) Load:
200 kPa
The average settlement is reduced with increasing the numberof piles. Arrangement 3 shows satisfactory performance forboth loads and beyond that the reduction procedure continued
at a slow rate. The average settlement due to uniform loading(solid line) is less than the average settlement caused byconcentrated loading (dashed line) in most cases. Because in
the piled raft under uniform loading there are regions without
loading and the settlement in this regions is not considerable.
Therefore, the resultant of average settlement in this case incomparison with the concentrated loading becomes less.
Another point that realized from Fig.4 is that increasing in raft
thickness may lead to reduction of average settlement, but
this reduction is not remarkable and with the increase in load
level the lines are approaching. The raft thickness of 1 metershows acceptable performance between different thicknesses.
Concentrated loading has unfavorable effect that is obvious indifferential settlement. Fig.5 shows the variation of theFig5. Variation of normalized differential settlement with thesquare root of the number of piles (a) Load: 150 kPa (b) Load:
200 kPa
For piled raft under uniform loading a decrease of normalized
differential settlement with increasing load level can beobserved. This is due to the plastic deformation under theedge being larger than those under center of the raft.
According to this figure, at arrangement 4, foundation shows
good performance. After the arrangement 4 (n p =81),
reduction procedure continued at a very slow rate andmodifying the pile number and arrangement is not required.
As previously mentioned by Horikoshi and Randolph [14],raft thickness describes the piled raft foundation stiffness
R

82

R

when raft width and soil configurations are constant. Thus, it
can be mentioned that the differential settlement increases
with decreasing foundation stiffness. Using thin raft (t r =0.5)
is less favorable, because it makes two or three times further
differential settlement than other conditions. On the other
hand, Using thick raft (t r =2) just to decrease the differential
settlement is not an economic plan. Raft thickness of 1 metershows satisfactory performance except at one case(concentrated load: 200 kPa).
Another unfavorable aspect of using thick raft is exceeding
maximum raft moment from the acceptable limits in design
procedure. Fig.6 clearly indicates rapid increase in maximum
raft bending moment when using larger value of raft
thickness.

admitted. However, the limit of 70 percent of applied load asan average seems good. In this case, arrangements 2 and 3 atraft thickness of 1 meter shows reasonable design alternative.

Fig7. Pile load ratio for different raft thicknesses (a) Load:
150 kPa (b) Load: 200 kPa
Therefore, assumption of two mentioned conventional
method about giving all loads to raft or to piles is not
acceptable and each of these elements play different role in
the foundation behaviour.
7. AN EXAMPLE OF OPTIMIZED DESIGNIn Table IV two arrangement of the parametric study iscompared with an example of optimized design for t r = 1 m
and load = 150 kPa. As suggested by Reul and Randolph [14],the piles under the core 25% of the raft area have the samelength (40 m) and others constructed shorter. In this example
other piles have 20 m length. According to the table,
reduction of the total pile length up to 30 percent makes little
increase in the average and differential settlements. Also, themagnitude of maximum absolute bending moment of the raft
becomes little larger and in some cases decreases incomparison with the studied ones. Therefore, in the optimum
design such cases should be considered to achieve the best
design alternative.

Fig6. Normalized maximum bending moment in raft fordifferent raft thicknesses (a) Load: 150 kPa (b) Load: 200 kPa
As mentioned above, raft thickness of 1 meter in this case isadmitted. Load type and magnitude has not had any
considerable on the normalized bending moment and bothtypes can be placed at a certain limit. At t r =1 arrangement 5
shows the best performance and indicates that using uniform
distributed arrangement of piles may cause reduction in raftbending moment and so the optimized design can beachieved.
Fig.7 shows the ratio of load taken by the piles. In someinstances piles take more than 90 percent of applied load. In
the scope of acceptable and economic design it is not
83

Table IV. Comparison of the foundation behaviour by equal
piles and combined piles∆w c-eS avg
Raft moment(cm)(kN.m)(cm)
R

Arrange3(Uniform)

Arrange3(concentrated)

Arrange5(Uniform)

Arrange5(concentrated)

Equal piles
Combined piles
Equal piles
Combined piles
Equal piles
Combined piles
Equal piles
Combined piles

4.4
6.17
4
6.2
3.7
5.6
3.8
5.6

1.2
2
3.8
3.9
1
1.8
1.3
2.5

[9] Poulos HG, “An Approximate Numerical Analysis ofPile-raft Interaction,”Int. Jl. for Numerical and Analytical
Methods in Geomechanics, Vol. 20, 1994, pp. 57-72.

R[10] Russo G, “Numerical Analysis of Piled rafts, ” Int Jl. for

2435
1610
1855
2625
1500
1690
1650
1670

Numerical and Analytical Methods in Geomechanics,
Vol. 22, No. 6, 1998, pp. 477-493.

[11] Brown PT and Wiesner T J, “The Behaviour ofUniformly Loaded Piled Strip Footings,” Soils andFoundations, Vol. 15, No. 4, 1975, pp. 13-21.
[12] Poulos HG, “Piled Rafts in Swelling or Consolidating
Soils,” Jl. Geotech.Engng, ASCE, Vol. 119, No. 2, 1993,
pp. 374-380.

8. CONCLUSION
In this paper a parametric studies on the piled raft foundation
behaviour has been done by accurate analysis and design
method (FEM) and it is understood that two mentioned
conventional design method are conservative. Because it has
been seen that for some cases, pile groups undergo just 40 or
50 percent of applied load and installation of the acceptable
raft above that can improve the design process. On the other
hand, using of piles as settlement reducer is not an economic
option and the design of raft in order to carry all
superstructure loads makes difficulties in this element
performance like exceeding bending moment. Therefore,
using more accurate analysis and design procedures help us toconsider different aspects of design and lead to choice of thebest option.

[13] Liu W and Novak M, “Soil-pile-cap Static Interaction
Analysis by Finite and Infinite Elements,” Canadian
Geotechnical Jl., Vol. 28, 1991, pp. 771-783.
[14] Reul O and Randolph MF, “Design Strategies for Piled
Rafts Subjected to Nonuniform Vertical Loading”, Jl. of
Geotec and GeoEnv Engng, Vol. 130, 2004, pp. 1-10.
[15] Oh EYN, Huang M, Surarak C, Adamec R and
Balasurbamaniam AS, “Finite Element Modeling for
Piled raft Foundation in Sand”, EASEC-11, Taiwan, 2008.
[16] Mandolini A and Viggiani C, “Settlement of Piled
Foundations,” Géotechnique, Vol. 47, 1997, pp. 791-816.
[17] Randolph MF, “Design methods for pile groups andpiled rafts,” state-of-the-art report. Proc. 13th Int. Conf.
Soil Mech. Found. Engng, New Delhi 5, 1994, pp. 61-82.

9. REFRENCES[1] O'Neill MW, Caputo V, De Cock F, Hartikainen J. andMets M, “Case histories of pile-supported rafts,” Reportfor ISSMFE Technical Committee TC18. Houston, TX:
University of Houston, 1996.

[18] Poulos HG, “Piled raft foundations: design andapplications,” Geotechnique, Vol. 51, 2001, pp. 95-113.
[19] Brinkgreve RBJ, Broere W. Plaxis 3D Foundation,
Version 1.6. User manual, Delft University ofTechnology & plaxis bv, The Netherlands, 2006.

[2] Van Impe W F and Lungu I, “Methods of analysis of piled
raft foundations,” A Report Prepared on Behalf ofTechnical Committee TC18 on Piled Foundations, 2001.

[20] Jeon SS, Lee JH and Lee CJ, “Slip effect at the pile–soilinterface on dragload,” Computers and Geotechnics 2004
Vol. 31 pp.115–26.

[3] Zeevaert L, “Compensated friction-pile foundation to
reduce the settlement of buildings on highly compressible
volcanic clay of Mexico City,” Proc. 4th Int. Conf. Soil
Mech. Found. Engng, London 2, 1957.

[21] Lee J, Kim Y, Sangseom J, “Three-dimensional
analysis of bearing behavior of piled raft on soft clay,”
Computers and Geotechnics, Vol. 37, 2010, pp. 103-114.

[4] Davis EH and Poulos HG, “The analysis of piled raftsystems,” Aust. Geomech. J, vol. 2, 1972, pp. 21-27.

[22] De Sanctis L, Mandolini A. “Bearing capacity of piled
rafts on soft clay soils,” Jl of Geotec & Geoenv Engng(ASCE), Vol. 132(12), 2006, pp. 1600–10.

[5] Hooper JA, “Observations on the behaviour of a piled-raftfoundation on London Clay,” Proc. Inst. Civ. Engnrs 55,
No. 2, 1973, pp. 855-877.

[23] Horikoshi K, and Randolph MF, “A Contribution to the
Optimum Design of Piled Rafts, ” Géotechnique, Vol. 48,
No. 2, 1998, pp. 301-317.

[6] Burland JB, Broms BB and De Mello VFB, “Behaviourof foundations and structures,” Proc. 9th Int. Conf. Soil
Mech. Found. Engng, Tokyo 2, 1977, pp. 495-546.

[24] Prakoso W A and Kulhawy FH, “Contribution to Piled
Raft Foundation Design,” Jl. of Geotechnical and
GeoEnvironmental Engineering, Vol. 127, No. 1, 2001,
pp.17-24.

[7] Sommer H, Wittman P and Ripper P, “Piled raftfoundation of a tall building in Frankfurt Clay,” Proc.
11th Conf. Soil Mech. Found. Engng, San Francisco 4,
1985, pp. 2253-2257.
[8] Franke E, “Measurements beneath piled rafts,” ENPC
Conference, Paris, 1991, pp. 1-28.
84

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Experimental Study of Suction-Monitored CBR Test on Sand-Kaolin Clay
MixturePurwana, Y.M., Nikraz H., and Jitsangiam P.
Curtin University, Australia
ABSTRACT
Conventional laboratory CBR test has been widely used for predicting bearing capacity of subgrade layer for pavement
design. In unsaturated soil, suction is one of the key parameters for understanding the soil behavior. The analysis of
CBR is commonly presented in CBR-water content relation. The information of CBR based on soil suction is very rare
and more study is still required. This paper presents a laboratory experiment of CBR test with direct suction
measurement. Suction-monitored CBR test is introduced by attaching tensiometers on CBR mold and its surcharge. The
standard compacted test on various proportions of sand-kaolin clay mixtures starting from 0% (pure sand), 5%, 10%,and 20% of clay were used. The tests were performed with different value of water content in both soaked and unsoaked
conditions. The results indicated that the CBR versus matric suction forms a bi-linear curve. The discussion is presented
in term of CBR-water content and CBR-matric suction relation.
Keywords: Bi-linear curve, CBR, matric suction, sand-kaolin mixture
was believed as the effect of compaction on the soil structureof the soil.
Black [4] carried out the study to estimate the CBR value ofcohesive soil indirectly using the soil plasticity and its
moisture content. Firstly, soil suction and true angle offriction were predicted using soil suction-water content andtrue angle of friction-plasticity index curves respectively.
From this, the predicted suction and the true angle of friction
were used to calculate bearing capacity, and finally the CBRwas estimated using the calculated bearing capacity.
Paraire [5] carried out the study for predicting CBR in
unsaturated condition indirectly using bearing capacity
approach. He performed the bearing capacity test using asmall cone with an angle of 30o and 2.03 cm in diameter
penetrated on the surface of silt and quartz sand specimen
with similar dimension of CBR mould. During the test,
suction was generated using suction plate apparatus and watercontent was also measured. The CBR was indirectly obtained
from the large number of suction-bearing capacity data. The
result indicated that there was a non-linear suction-CBR
relation in which the increase in suction caused the increase in
CBR.
The non-linearity of CBR with respect to matric suction was
also resulted from the study carried out by Sivakumar andTan [6]. They investigated the effect of compacted watercontent on CBR of tills, while matric suction was taken into
consideration using a pressure plate. The effect of matricsuction was presented indirectly by water content, andindicated that water content has a significant effect on CBR.
The result also indicated that the rate increase in CBR at wet
side of optimum water content was lower than that at dry side.
The effect of matric suction on CBR value has also beenstudied by Ampadu [7] on decomposed granite. He focusedon the effect of drying and wetting of remolded specimen on

1. INTRODUCTION
The California Bearing Ratio (CBR) has been used as a semi
empirical approach for predicting bearing capacity of
subgrade for pavement design. The method was firstly used in
1920’s in the California State Highway Department and wasadapted by US Corp of Engineer in the 1940’s for military
airfield. After the Second World War, the CBR was also usedin UK and has spread to European countries [1]-[2]. Due to its
simplicity and relatively low cost, this method has been
widely used around the world for flexible pavement design.
The CBR test is performed by penetrating a plunger withconstant penetration rate of 1.27 in/min on compacted
specimen on rigid mould with diameter of 152.4 mm and aheight of 177.8 mm as describe in ASTM D 1883-07. Thereare 2 types of CBR test associated with the treatment of thesample: unsoaked and soaked. The unsoaked CBR test is
performed to simulate the natural condition whereas thesoaked one is performed to replicate the worst condition suchas high water table in the field, in which the soil layer isconsidered to be in saturated or nearly saturated condition.
The intensive study of the CBR and its relationship withwater content and dry density has been studied by Davis [3]using a relatively wide range variation of soils from heavy
clay to sandy gravel. The relation was presented in semi
logarithm diagram, showing that the log CBR value wasaffected by water content, dry density and structure of soil
particle. As water content increased, the CBR value went
down linearly for plastic soil, whereas for non-plastic soils
the relation was non-linear. The non-plastic the soil exhibited
increasing in CBR at initial range of water content, anddecreasing as the water content increased. This non linearity

85

described in ASTM D 698. The grain size distribution curvesof all mixtures are shown in Fig. 2.

CBR value. The result indicated that from the OMC the CBRtends to increase rapidly as water content decreased. The
CBR was drop as water content increased due to soaking. In
his study, matric suction was obtained indirectly from SWCC
using filter paper method.
The aforementioned studies indicated that many moreinvestigations regarding the effect of matric suction on CBR
were still required. This paper presents the experimental
study of CBR on various proportion of sand-kaolin mixtureswith direct measurement of matric suction using tensiometer.
2. SUCTION-MONITORED CBR DEVICE
The suction-monitored CBR is the device for CBR test inwhich matric suction is monitored during the test. It is madeby a modification of conventional CBR by attaching 1-bar
miniature tensiometers on its mold and surcharge weight. The
set-up and configuration of the test is shown in Fig. 1. The
tensiometers were connected to the data logger or readout unit
so that the suction measurement together with the loading anddisplacement can be recorded continuously during the test.

Surcharge

Fig. 2. Grain size distribution curves

Tensiometer3 cm

Specimen

6 cm

3 cm

Fig. 3. Compaction curves

Fig. 1: The setup of modified CBR test (not to scale)Fig. 3 and Table 1 show the compaction curves andcompaction characteristic respectively. During compaction,
the small particles of kaolin clay fill the voids causing theincrease in MDD and the decrease in void ratio. This result
has a good agreement with the result of Mullins andPanayiotopoulos [8] and Chiu and Shackelford [9]. The
optimum water content (OWC) and maximum dry density(MDD) were two referenced parameters for the compactionof CBR specimens. The relatively high water content in 100%sand was caused by the high portion of fine size of sand (>50%).
Table 1. Compaction characteristic of the sand-kaolin mixture
Sand-kaolin
MDD
OWC
Void ratio
clay mixture(kg/m3)(%)(e)
100% sand
1700
13
0.55
95:5
1830
11
0.45
90:10
1970
9.8
0.35
80:20
2030
9.1
0.31

3. TESTING PROGRAM
3.1 Preparation
Sand-kaolin mixture with different proportion of 100% sand,
95:5, 90:10, 80:20 dry weight were investigated. The sandwas taken from Baldivis sand pit, Western Australia, and has
the specific gravity of 2.63. Grain size analysis indicated thatit consists of 99.74 % fine and medium size of sand and 0.26
% of fines. According to the ASTM D 2487, this material isclassified as poorly-graded sand (SP). The trade mark ofkaolin clay was obtained from the UNIMIN Australia Ltd.,and has the liquid limit and plasticity index of 48.1 and 17.4
respectively. The dry sand and kaolin clay were then mixed
with the pure water according to the desired proportion andput in a sealed plastic bag for at least 72 hours for curing andwater content equilibration. The compaction test was
performed following the Proctor compaction procedure as86Prior to the CBR compaction, the orifices prepared for
attaching the tensiometers on the mold were blocked tightly
with electrical tape. This was done to avoid the specimen
coming out from the mold through these holes during
compaction. The CBR tests were performed on both“saturated” and unsaturated conditions, represented bysoaked and unsoaked specimen respectively.
It is essential to ensure that before and during matric suction
measurement, the tensiometers were completely free of air
bubbles. The presence even a small air bubble would haveaffected the performance of the tensiometer by increasing the
response time. The saturation of tensiometers was thenrequired. It was performed by immersing its ceramic disk andreservoir in the distilled water and then was vacuumed using a
desiccator and a vacuum pump for around 2-3 hours.

Fig. 4. The effect of kaolin clay content on CBR
4.2 The Effect of Water Content and Matric Suction on
CBR
Fig. 5 quantifies the effect of the decrease in water content
due to air-drying on CBR. The CBR-water content relation
forms a non-linear curve. Until its water content reaches the
OWC, the CBR increases slightly as the water content
decreases, and a significant increase rate in CBR occurs when
the water content is less than OWC. The result has a good
agreement with the result of Ampadu [6]. In general, thelonger the air-drying period, the lower the specimen watercontent and the higher the CBR value.

3.2 Testing Program
Prior to plunger penetration, the tensiometers were installed
through the orifices on the CBR mold and its surcharge insuch a way that its high air entry disk (HAED) of tensiometers
has a good contact with the specimen. High quality
removable adhesive and clamping set were utilized to holdthe tensiometers in their position. Pneumatic machine wasutilized to penetrate the CBR plunger on the specimen with aconstant rate of 1 mm/min. A load-penetration curve
produced from the test was utilized for determining the CBRvalue. By definition, the CBR value is the magnitude of stressat which the specimen has been penetrated at 1 in (2.54 mm)compared to a standard stress of 6.9 MPa. During the test, thematric suction of the bottom and upper layers of the specimen
were recorded. After the CBR test, the water content of thebottom, middle and top layer of specimen were determined.
In this study, various water contents were obtained byair-drying for different time periods from unsoaked
specimens.

OWC

4. RESULTS
4.1 The Effect of Clay Content and Air Drying Period on
CBR
Fig. 4 shows the effect of kaolin clay content on CBR for
various periods of time. The CBR of 100% sand (0% kaolin
clay content) are not significantly affected by the watercontent, as the CBR values remain around 15-20% at any
water content. The increase of kaolin clay content up to 5%
causes the decrease in CBR in both soaked and unsoaked
condition, but it causes a slight increase in CBR when thespecimens were dried up for 1 or 2.5 day. The significant
increase in CBR is shown by 10% kaolin clay content (90:10
mixture) when the specimen was dried up. The mixture of
90:10 exhibits the best proportion in term of CBR (eventhough according to compaction curves, the highest MDD
was resulted from 80:20 mixture). The CBR of 90:10 mixture
increases from 18.6% in soaked condition to 28.3% in
unsoaked condition. Again, the CBR increases to 41.3% and48.5%, when the sample was dried up for 1 and 2.5 days
respectively. Beyond 10% kaolin clay content, the CBRdecreases even though the samples were air-dried.
.

Fig. 5. The effect of water content on CBRThe effect of water content and matric suction on the strengthof compacted sand-kaolin mixtures has been investigated aspart of the author’s study. The result indicated that to someextent the presence of kaolin clay affected the shear strength
through the increase of effective cohesion and the decrease ofeffective internal friction angle. However, the cohesion itself
was affected by the amount of water in the mixture.
Accordingly, it may be deduced that the CBR of compacted
sand-kaolin clay mixture is affected by the combination effectof kaolin clay and water content altogether.
It has been recognized that the decrease in water content leads
to the increase in matric suction. The effect of the increased
matric suction on the CBR is presented in Fig. 6. This result87may be explained as the decrease in water content due todrying causes the increase in water tension amongst the soil
particles causing the increase in matric suction, and as aconsequence, causing the increase in strength.
It can be observed from the figure that the relation between
CBR and matric suction forms a bi-linear curve. The first partof the curve starts from near zero matric suction (soaked
sample) to 5-7 kPa, at which the CBR increases sharply, andthe second part starts from 5-7 kPa with a relatively slower
increase in CBR. It also can be seen that the matric suction
has never attained zero even though the sample was soakedfor 4 days. The soaked CBR test was not performed in the
fully saturated condition. The reason is that the test was
performed 15 minutes after removing the mold from the bath.
Some of the water in the specimen may infiltrate through thebase, and some of them may evaporate.

sharply is started at near zero matric suction (soaked
sample) until 5-7 kPa, continued by a relatively slower
increase in CBR at the second part of the curve. For
100% sand, the second part of the curve exhibits thedecrease in CBR even though the matric suction isincreased.
6. Further study is required using the device with higher
matric suction capacity such as suction probe, and alsousing the wider range of soil types.
6. ACKNOWLEDGEMENTS
The first author gratefully acknowledges the Indonesian
Ministry of Education for providing funds through the DIKTI
Scholarship Project.
7. REFERENCES[1] Croney, D. and Croney, P., The Design and Performanceof Road Pavements, 2nd edition, Mc-Graw Hill Book
Company, 1991, Part 3, pp. 159.
[2] Asworth, R. ‘Highway Engineering’, Heinemann
Education Books, London, 1972, Chapter 11, pp. 190.
[3] Davis, E.H. 'The California Bearing Ratio Method forthe Design of Flexible Roads and Runway',
Geotechnique , Vol. 1, No. 4, 1946, pp. 249-263.
[4] Black, W.P.M. 'A Method of Estimating the California
Bearing Ratio of Cohesive Soils from Plasticity Data',
Geotechnique, Vol. 12, No. 4, 1962, 271-282.
[5] Paraire J., Suction Tests on CBR-Diameter Specimens.
The Bearing Capacity-Suction Relation', Transport andRoad Research Lab. (TRRL), 1987.
[6] Sivakumar, V. and W.C Tan, W.C., 'CBR, UndrainedStrength and Yielding Characteristics of Compacted
Tills', Unsaturated Soils: Proc. of the Third International
Conference on Unsaturated Soils, 2002, pp. 657-661.
[7] Ampadu, S., 'A Laboratory Investigation into the Effectof Water Content on the CBR of a Subgrade Soil',
Experimental Unsaturated Soil Mechanics, Springer
Berlin Heidelberg, 2007, pp. 137-144.
[8] Mullins, C.E. and Panayiotopoulos, K.P., ‘Compactionand Shrinkage of sands and sands-kaolin mixture’, Soiland Tillage Research, 4, 1984, pp. 191-198.
[9] Chiu, T.F. and Shackelford, C.D., ‘Unsaturated
Hydraulic Conductivity of Compacted Sand-Kaolin
Mixture, J. of Geotechnical and Geoenvironmental
Engineering, Vol. 124, No.2, 1998, pp. 160-170.

Fig. 6. Bi-linear curve of CBR versus matric suction relation
5. SUMMARY AND CONCLUSION
Based on the laboratory testing on CBR with suction
measurement, it can be concluded:
1. Suction-monitored CBR is modification of conventional
CBR by attaching tensiometers on its mould andsurcharge weight.
2. The CBR test was performed on sand-kaolin clay
mixtures of 100% sand, 95:5, 90:10, and 80:20 dry
weight proportion, combined with the variation of watercontent and matric suction using air-dying method.
3. The CBR value of each mixture was presented in respect
to water content and matric suction. The decrease inwater content leads to the increase in matric suction, andthe increase in matric suction leads to the increase in
CBR.
4. The relation between CBR versus water content and also
the CBR versus matric suction are non-linear. Until thewater content reaches the OWC, the CBR increases
slightly as the water content decreases. A significant
increase rate in CBR value occurs when the specimen
water content is less than its OWC.
5. In this study, the non-linearity of CBR in respect tomatric suction forms a bi-linear curve for all mixtures.
The first part of the curve at which the CBR increases
88Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051{The person who says it can not bedone should not interrupt the person
doing it-Chinese Proverb}

85

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Foundation Strengthening of Historic Buildings by Micropiles
Professor Dr. Azza M. Elleboudy
Benha University, Egypt
ABSTRACT
Cairo contains a treasure of historical buildings and monuments. Several magnificent monuments have deteriorated
over the years due to environmental changes. The restoration schemes of two important monuments are the subject of
this paper. Damage caused by rise of ground water table and a recent earthquake resulted in cracks in their walls,
facades, columns and arches. Restoration started with strengthening the bearing soil strata and the foundations
with micropiles to transfer a part of the loads to deep granular soil layer below the weak heterogeneous fill. They were
reinforced by steel pipes, and injected with cement grout under different pressures. Micropiles have provided not only
an excellent structural support but also a system that minimized settlement, and allowed their execution under the
existing restricted access conditions with minimal disturbance of the buildings during underpinning.
Keywords: Micropiles, Foundations, Monuments, Restoration
Coptic Christian church in Cairo. It was built in the 7th
century.

1 INTRODUCTION
Cairo and its environs preserved and contained a parade ofhistorical buildings and monuments famous for theirprecious value and fascinating design. Several Islamic andChristian monuments have suffered from severe
environmental attack over the years and showed serious
signs of distress. Structural cracks caused by differential
settlement of the foundations were noted in walls, facades,
columns and arches of these invaluable buildings. Thesephenomena called for a comprehensive geotechnicalinvestigation to outline the causes of the problem andsuggest the proper strengthening and rehabilitation schemes.
Micropiles were a rather successful means in the restoration
process of these important structures [1].

2 MICROPILES
Micropiles are drilled, cast-in-place, and grouted piles withsmall diameter, normally between 125 and 250 mm. They
have the capability of sustaining high loads. The main
advantage of employing micropiles in the renovation of thehistoric buildings is their suitability for headroom restricted
access, their suitability for wide range of soil conditions andtheir ability to penetrate obstructions. Furthermore,
micropiles can be constructed with minimal noise, vibrationand disturbance of the structure and the surrounding
buildings, at any angle below horizontal.
Micropiles are mainly used for:a. Foundation rehabilitation with underpinning, upgrading
foundation capacity, and seizing the foundation movement
b. Soil reinforcement as in landslide stabilization, soil
strengthening, and settlement reduction
c. Resisting dynamic loads
d. Excavation support
e. Resisting uplift forces
Micropiles are installed mainly by two methods;
Replacement and Displacement methods. Piles, when driven,are defined as “Displacement Piles” because their installation
methods laterally displace the soils through which they are
installed. The most commonly used in ancient buildings are
the “Replacement Piles” which are formed by drilling andgrouting. The load is mainly carried by the micropiles
reinforcement and transferred through the grout to thesurrounding soils by high values of interfacial friction. The
majority of micropiles are between 100 and 250 mm indiameter, 15 to 30 m in length and 250 to 1000 kN incompressive or tensile service load [9].

Micropiles were first installed in Italy in the 1950s and were
utilized as a method to underpin historic buildings andmonuments. In the 1970s, this new technique wasintroduced in the U.S. In the late 1980s, it gained a wide
acceptance as a means for strengthening and underpinningfoundations of ancient buildings and existing structures aswell in the construction of new civil projects in many
countries. Micropiles were successfully used in Turkey [2],Spain [3], France [4], Japan [5], Canada [6] and Australia[7] as well as many other countries around the world.
Micropiles were used in Egypt for the first time in 1993 in therestoration of Sultan El-Ghoury Mosque in Cairo. Since then,they have been utilized in the rehabilitation of many historic
structures [8] such as Al-Azhar Mosque, and the Hanging
Church, the subject of this paper. Al-Azhar Mosque is one of
the most important mosques in Cairo, built in the Fatimid
era in year 970. The Hanging Church is the most famous91

It mostly yielded very low computed capacities compared tothe observed capacities, showing that this method is not
appropriate for compaction grouted piles.

2.1 Design Aspects
Several design procedures were suggested by many
researchers [10]. The design methods are intended to provide
an axial capacity for initial design which should be checkedwith results of load tests. Locally, the methods widelyadopted are the Empirical Methods described as follows.
2.1.1 Anchor System
Based on field experience [11], the design method foranchors grouted under pressures of less than 1.0 MPa can beused to calculate the capacity of the micropile, Q u , obtainedby side friction.
Q u = L x n x t a n φ ’ (kN)(1)
where,
L : pile bond length in meter
φ ’ : effective angle of shearing resistancen : factor in kN/m, affected by the drilling technique, depthof overburden pressure in the range 30 to 1000 kPa, insitu stress field, and dilation characteristics of the soil.
Table I gives the values of “n” for a given soil type based onfield experience. They were obtained for normallyconsolidated materials for borehole anchor diameters ofapproximately 0.1 m. If the drilling diameter changes
significantly, n should be proportionally modified.

2.2 Method of Construction
2.2.1 Placing the casing into the groundTo place the casing into the ground, drilling is used because
the placing can be done in almost every ground conditions. If
the casing is drilled properly, it does not cause significant
disturbance to adjacent structures or to the structure being
underpinned. Where, a casing is rotated with a "drilling bit"at the bottom to cut the soil and fluid circulates through thecasing to flush the soil cuttings through the space between thecasing and the ground.
2.2.2 Placing the steel reinforcement
Reinforcement is placed in the casing. It can be a single
rebar, cage of reinforcement bars or a steel pipe. If thereinforcement is a steel pipe, it can be provided with sleevesand packers. The grouting can be performed through the steel
pipe at specific horizons. If the reinforcement is not a pipe,
either a plastic pipe for grouting may be placed near thereinforcement or grouting is done directly through the casing.
2.2.3 Injecting the grout
The typical grout used in micropile practice is a mixture ofwater and cement with a water/cement ratio between 0.40
and 0.55. Sand is sometimes added to decrease the cost of the
mixture.
In many cases and before injecting the grout, the casing isfilled with grout placed by tremie to push the drilling fluidout of the casing. Reinforcement is sometimes installed after
tremie grouting. The subsequent pressure injection of thegrout can be done in different ways; through the casing as it isremoved, or through the steel pipe used for reinforcement, or
through a plastic pipe placed near the reinforcement.
Repeated pressure grouting is used in some cases. It costs
more than the single injection, but the capacity of the
micropile increases due to the better grout-soil bondobtained. Typical dimensions and grouting pressures
reported by various authors [9], [13] and [14] from thestate-of-the-practice are summarized in Table IV.

Table I. Values of factor n (kN/m)
Type of Soil

Permeability (m/sec)n

Coarse sand &Gravel> l 0 -4

400 to 600

Medium to fine sand10- 4 to 10- 6

130 to 165

2.1.2 Lizzi’s Method
Lizzi proposed a simple empirical formula to evaluate theultimate load of the micropile [12], which is:

Q ult. = D x L x K x I(2)
Where,
D : nominal diameter of the micropile in cm (drilling
diameter)
L : length of the pile in cm
K : pile-soil adherence coefficient, Table II
I : dimensionless coefficient depends on the micropile
nominal diameter (drilling diameter), Table III
Table II. Values of K

Table IV. Typical characteristics of a micropile

Table III. Values of I

BoringReference Diameter

Length[9](mm)100-250(m)
Up to 40

Grouting
Pressure

Soil TypeK (kg/cm )Pile Diameter

I

Soft soil
Loose soil
Soil of averagecompactness
Very stiff soil,
gravels, sands

0.5
1.0
1.5

10 cm

1.00

[13]

100-250

20-30

(MPa)
Low: 0.2-0.5
High:1.0-2.0
Up to 8 MPa

15 cm

0.90

[14]

67-89

4-4.9

Up to 9 MPa

20 cm

0.85

2.0

25 cm

0.80

2

ServiceLoad150-600 kN
300-1000 kN
Ultimate99-140 kips

3 AL-AZHAR MOSQUE
It is one of the most important mosques in Cairo, Egypt (Fig.
1). Al-MuizzliDinAllah,the FatimidCaliphate commissioned
its construction for the newly established capital city in year
970. After its dedication in 972, and with the hiring by

The empirical design method proposed by Lizzi, developed
basically for root piles with relatively low grouting pressures.
92mosque authorities of 35 scholars in 989, the mosque slowly
developed into what is today the second oldest continuously
run university in the world, Al-Azhar University, which has
long been regarded as the foremost institution in the Islamic
world [15].

Fig. 2. Al-Azhar Mihrab

Fig. 1. Al-Azhar Mosque interior yard
The Mosque has drawn the attention to its precious value andfascinating design features with their clever and inspired
craftsmanship (Fig. 2 and 3). Over the course of amillennium-long history, the mosque has been alternately
neglected and highly regarded by Egypt’s rulers, andnumerous expansions and renovations took place. Recently,
Egypt showed differing degrees of deference to the mosque.
Today, Al-Azhar remains a deeply influential institution in
Egyptian society and a symbol of Islamic Egypt.

Fig. 2. Artistic details in the ceiling
3.1 The Problem
The mosque recent area is about 12000m2. It has 3 gates, 5
minarets, and 380 columns. It included several parts built indifferent eras starting with the Fatimid epoch in 972 to
Mamluki and Othmani periods. In the recent time, whichstarts in 1936, the main façade was built in 1970. Therefore,
every edition was built on a different foundation at various
foundation level, size and material.
The walls, arches and footings of Al-Azhar Mosque weremade of masonry (bricks & stones) and clay-lime mortar. Its
walls and columns were supported by strip footings built onheterogeneous fill that extended to a considerable depth.
After the 1992 strong earthquake that hit Cairo, wide anddeep cracks spread in the arches and walls of the mosque, andinclination was noticed in many columns.
3.2 Rehabilitation Plan
An engineering committee including geotechnical andstructural engineers, as well as architects studied these
phenomena to outline the causes. A restoration andrenovation program was initiated. It started with a foundation
investigation. Test pits were executed to inspect thefoundations. They revealed that the bearing walls were
resting on weak lime stone footings 25cm wider than thewall, at depth varied from 1,40m to 2.40m below the floor
level, and at depth of 0.90m under the columns. The grade93

beams, when existed, were built with old red bricks where themortar consisted of lime and clay, and they were muchdeteriorated. In some places, there were no foundations undersome walls and columns which aggravated the problem. The
deep borings executed at the site indicated that the soil isgenerally formed of a heterogeneous fill layer started fromthe mosque floor to a depth varied between 3 and 7m
followed by a silty clay layer contained limestone fragments
to a depth around 9m. The bottom layer consisted of gradedsand with traces of silt and clay.
The geotechnical engineers recommended the use of
micropiles to reinforce the existing foundation. The footings
were underpinned using micropiles 15 cm in diameter,
12-15m in depth, reinforcement was high tensile steel pipe of
9 cm outer diameter and 1.0 cm thick. Grouting was carriedout using cement grout under pressure (low pressure of 0.2 0.5 MPa & high pressure of 1-2 MPa) [8].
Micropiles were connected together by a reinforced concrete
continuous cap. Caps were built on the sides of walls orcolumns at the same level under the floor. The cap wasconnected to the masonry strip footing by means of shear
connectors. Figure 4 illustrates the method of underpinningthe masonry footings supporting columns and walls.
Parts of the walls and columns located upper the level of pile
cap must be strong or strengthened by grouting. The masonrywalls and columns were repaired by jet grouting under low
pressure of 0.1 MPa.

4.1 Historical Background
The Hanging Church was built in the 7th century, probably
on the site of a 3rd or 4th century church for the soldiers of
the bastion. It has been rebuilt several times since then,including a major rebuild under Patriarch Abraham in the10th century. By the 11th century, it became the official
residence of the Coptic patriarchs of Alexandria and several
Coptic synods were held in the church. The main furnishings
and the pulpit and screens date from the 13th century. The
church was largely rebuilt by the Pope Abraham (975-78)
and has seen many restorations over the years including thevery recent one finished in 2009.
4.2 Building Description
Since the Hanging Church stands with its Basilica-stylearchitecture on top of the Roman Fortress of Babylon in Old
Cairo, its nave is suspended over a passage. The land surface
has risen by some 6 meters since the Roman period so that the
Roman tower is mostly buried below ground, reducing thevisual impact of the church's elevated position.

Fig. 4. Underpinning foundations with micropiles

Fig. 5. The Hanging Church entrance

4 THE HANGING CHURCH

The entrance to the Hanging Church is via a beautifully
decorated gate under a pointed stone arch. The nineteenth
century facade with twin bell towers is then seen beyond anarrow courtyard decorated with modern art biblical designs.
This leads into an open courtyard, flanked by mosaics, fromwhich there are 29 steps to the church. At the top of the stairs
are three wooden doors decorated with geometric patterns,
framed with decorative carvings in the stone wall (Fig. 5). Up
the steps and through the entrance is a further small courtyard
leading to the eleventh century outer porch. Inside, only the

The Hanging Church is the most famous Coptic Christianchurch in Cairo (Fig.5). The Saint Virgin Mary's Coptic
Orthodox Church which known as the Hanging or the
Suspended Church is one of the oldest churches in Egypt. It is
located in a religious compound in Old Cairo, adjacent to the
Amr Ibn Al-Ass Mosque, Ben Ezra Synagogue and acollection of churches. The Hanging is named for its location
above a gatehouse of Babylon Fortress [15].

94section to the right of the sanctuary, above the southernbastion, is considered original. Nevertheless, the Hanging
Church remains one of the most impressive churches in thecity, remarkable for its marble pulpit, inlaid screens, icons
and murals. The timber roof of the nave recalls Noah's Ark(Fig. 7).

In 1997 the Supreme Council of Antiquities (SCA) launched
a comprehensive restoration project to preserve Egypt’s
Coptic shrine and return such a distinguished church to its
original splendor. The restoration project of the Hanging
Church in Old Cairo was started in 1997 and completed in2009. The restoration process was indeed very complex
because it is a comprehensive project that deals with
underground water; the most pressing problem facing Islamic
and Coptic antiquities since 120 years, besides thereinforcement and repair of most structural elements andarchitectural features of the monuments.
There were many important parameters and considerations inthe selection of the appropriate restoration schemes. Saving
the monument should start with preventing the main cause of
the problem. It was important to lower the ground water table
and stop the leaking water. Monitoring the water table in thisarea called for an overall dewatering project to be
implemented in old Cairo without causing any excessive
settlement to the existing buildings.
The aim of restoration, which was carried out in phases, wasmainly to reduce the water leakage into the church andstrengthen the church’s foundations and the Babylon fortressin order to protect them from future settlements and relative
movement that causes further damage.
4.4.1 Lowering the groundwater level
The first stage included reducing the level of the water in thefortress. Deep wells were dug in the area around the fortress
to trap any excess water. Two additional wells were dug infront of the fortress, and in front of the newly revealed door
to the adjacent Mosque of Amr. The water gathered in thewells was siphoned out via pipes to the main sewage system.
Then, a comprehensive dewatering project was carried out inthe whole area of old Cairo to lower the ground water table
and stabilize it at a reasonable depth below the foundations of
the Church and the surrounding monuments.
4.4.2 Strengthening the foundations
A geotechnical investigation was carried out at the church
site which revealed that soil layers are mainly composed ofthick layer of heterogeneous fill that extended to 8m below
the ground surface in some spots at the site. This layer was
followed by sand interlayered with silty clay and extending atleast to a depth of 10m; the end of borings. The walls wererested on deteriorated lime stone strip footings. The proposed
foundation rehabilitation was to transfer at least half of theloads carried by the masonry bearing walls to a deeper
stronger subsoil layer by means of micropiles. The
micropiles were the most suitable means for foundation
reinforcement to ensure the least disturbance for the oldmonument and to manage its installation in locations withlow headroom, and narrow and restricted access. The loadtransfer from the old footings to the piles were made byconstructing reinforced concrete beams as pile caps,
transversally connected to the existing footings by sheer
connectors as illustrated in Fig. 8. Following that there-enhancement of the old masonry footings took place.

Fig. 6. Detail of decoration over the door-arch

Fig.7. Hanging Church barrel-vaulted aisle-ceiling
4.3 Damage Assessment
The Hanging Church, like other monuments located inheavily populated areas, was suffering from environmental
hazards including air pollution, a high subsoil water level, ahigh rate of humidity, leakage of water from the outdated anddecayed sewage system installed 100 years ago.
The environmental impact on this magnificent monument
resulted in a serious static crisis. The main reason for suchdamage and deterioration is attributed to human abuse. The
increase of the population and economic activities around thebuilding, rapid urbanization in the whole area, andconsequently, the rise of the ground water table led tohazardous environmental changes that damaged the integrityof the church [16]. Above that, the adverse effects of the
1992’s strong earthquake increased the number of cracks allover the church’s walls, barrel-vaulted aisle-wooden ceilingand door arches. Ground water had flooded the Babylon
fortress, which supports the church and leaked through to thechurch. The ground in front of the Hanging Church had
caved in. The church priests thought that the church was on
the brink of destruction.
4.4 Restoration Project

Limestone wall
95
φ16mm

gained a wide acceptance as a means for strengthening andunderpinning foundations of ancient buildings and existing
structures as well as construction of new civil projects inmany countries.
Since1993 up to 2001, micropiles have been used tostrengthen 12 Egyptian historic buildings, 11 of them are in
Cairo. These micropiles were mostly vertical with working
load ranged between 15 to 45 tons and reached to 80 tons in
few cases. Diameters of micropiles were between 14 to 20
cm and their lengths ranged between 12 to 23.5 m. The outer
steel diameters of micropiles varied between 88.9 to
127 m m , a n d both high and low pressures were used intheir grouting.
This paper discussed the use of micropiles to strengthen the
bearing soils and the foundations of two significant historicbuildings in Cairo, Egypt, namely, Al-Azhar Mosque and theHanging Church. Both of them are masonry structures withbearing walls founded on continuous stone brick footings
supported by heterogeneous fill. The recent severe
deterioration was due to rise of water table and the strong
1992 earthquake. Micropiles were the essential factor in the
rehabilitation and restoration process. They weresuccessfully installed to transfer a good part of the load todeeper subsoil and to strengthen the foundations. Then, awhole renovation of all the structural members, architecturalfeatures and artistic details took place to reserve themonuments’ glory and reopen them to worshippers andtourists.

Fig. 8. Construction detail of micropiles
4.4.3 Repairing the super structural elements
The walls and the columns that hold the church also needed
reinforcement as well as ceilings of the church. Injection of
the church supporting walls with a filler material of the same
type used in the original construction, to replace the original
filler materials lost as a result of the variance in ground waterlevels. Repair of all wall cracks. Masonry stitching of thewalls of the Church was carried out. The walls were
reinforced, missing and decayed stones were replaced, andmasonry was cleaned and desalinated.
4.4.4 Dovetailing the stones of the Babylon Fortress
The walls and the ceiling of the fortress were restored as wellas the artwork throughout the fortress (Fig. 9).
4.4.5 Repair of all the fine detailed works
The decorations and icons of the church were also subjected
to fine restoration and new lighting and ventilation systems
have been installed. Every effort was made to ensure that all
the original architectural features were retained.
Moreover, the restoration work had been extended beyond
the church’s walls to reach its neighboring monuments andthe streets surrounding it, and the entire neighborhood was
being revived and upgraded.

6 REFERENCES[1] Elleboudy A and Hamdy G, “Strengthening and repair ofa monument in Cairo,” Proc. 1st European Workshop on
Structural Health Monitoring, Paris, 2002, pp.
1235-1242.
[2] Togrol E, “Ground improvement and piling,” Proc.
14th ICSMFE, Hamburg, Vol.2, 1997, pp. 907-910.
[3] Babu G et al., “B e a r i n g capacity improvement
using micropiles: a case study,” Civil Engineering,
ASCE, Vol. 73, No.10, 2004, pp. 92-99.
[4] Maleki K and Frank R, “an approach for the analysisof axially loaded micropile groups,” Proc. 14th
ICSMFE, Hamburg, Vol. 2, 1997, pp. 1107-1110.
[5] Kamon M, “Case studies of reinforced ground w i t hmicropiling and other improvement techniques,"Proc. Symp. on Prediction versus Performance inGeotechnical Engineering, Bangkok, Thailand, 1992,
pp. 115- 125.
[6] To P and Watts B, “Tension/compression load
testing of a minipile,” Proc.13lh Conf. ICSMFE, New
Delhi, India, Vol. 3, 1994, pp. 1219-1222.
[7] Makarchian M and Poulos 1I G, “Underpinning by
piles: a numerical study,” Proc. 13lh Conf. ICSMFE,
New Delhi, India, Vol. 4, 1994, pp. 1467-1470.
[8] Yusef A, “Foundation repair for Al-Azhar Mosque,”Technical Report, ACTI, Egypt, 1998, pp.1-25.
[9] Egyptian Code of Practice for Soil Mechanics andDesign and Construction of Foundations, Vol. 4, Deep
Foundations, 2005.

Fig. 9. Babylon Fortress
5 CONCLUSIONS
Micropiles have been utilized for underpinning of historicbuildings and monuments since they were successfully used
for this purpose in Italy during the seventies. This technique96

[10] Micropile Design and Construction Guidelines,
Implementation Manual. FHWA – SA-97-070, 2000.
[11] Schwarz H, et al., “ S p e c i a l use of micropiles andpermanent anchors,” J. of Geotechnical Engineering,
ASCE, Vol. 128, No. 11, 2003, pp. 945-957.
[12] Lizzi F, The Static Restoration of Monuments. Genoa,
Italy: Sagep Publisher, 1988.
[13] Bruce D, et al., “A primer on micropiles”, Civil
Engineering, ASCE, Vol. 65, No. 12, 1995,
pp.51-54.
[14] “Evaluation of Compaction Grouted Minipiles,” theNorthwestern University National GeotechnicalExperimentation Site, 1993, pp. 1-62.
[15] Koraim S, Cairo: A City of 5000 Years. Cairo:
Egyptian Book Authority, 1995.
[16] Elleboudy A and El-Sohby M, “Environmental effects
on the groundwater in Greater Cairo,” Proc. Int. Symp.
on
Environmental
Geotechnology,
Allentown,
Pennsylvania, USA. Vol. 1, 1995, pp. 167-180.

97

Second International Conference on Geotechnique, Construction Materials andEnvironment, Kuala Lumpur, Malaysia, Nov. 19-21, 2012, ISBN: 978-4-9905958-1-4Group Effects of Piles Due To Lateral Soil Movement
Hongyu Qin1 and Wei Dong Guo 2
School of Engineering, Griffith University, Gold Coast, QLD 4222 Australia
2
School of Civil, Mining and Environmental Engineering, University of Wollongong, NSW 2522, Australia
1

ABSTRACTLaboratory model tests have been conducted to investigate the responses of piles subjected to lateral soil movement.
The results of a single pile test and four tests on two piles arranged in a row perpendicular to the direction of soil
movement are presented. The development of maximum bending moment, maximum shear force, and pile deflection
with soil movement and the largest pile response profiles for the single pile and pile groups are compared. Group effectwas evaluated using group factor which is defined in terms of the measured maximum bending moment. The major
findings are (1) the pile head conditions (free or capped) are insignificant on piles subjected to lateral soil movement
when arranged in a row, (2) the group factor decreases as the pile spacing reduces, (3) a linear relationship exists
between the maximum bending moment and maximum shear force for both the pile groups and single pile.
Keywords: Piles, Group Effect, Lateral Soil Movement
soil movement. Four tests on a group of two piles in a row
with the pile head capped and a single pile of free head wereconducted. Typical pile responses are presented. Groupfactors are used to quantify the group effect and comparedwith previous experimental and numerical analysis results.

1. INTRODUCTION
Structures built on piles are vulnerable to lateral forces causedby soil movements, which may be seen when they are used inslope stabilization, to support bridge abutments, and as
foundations of tall buildings adjacent to tunnelling andexcavation or lateral spreading in liquefied sand during
earthquake. Lateral loads generated by soil movements
induce additional deflections and bending moments in piles
or pile groups, which may undermine the structural integrityof the piles or groups and leads to serviceability or even
failure of the piles. Although this characterization has been
extensively studied through centrifuge modeling and 1g small
scale model tests in [1]-[9], field tests in [10]-[11]andtheoretical and numerical analysis in [12]-[16], the pile soilinteraction mechanism is still not clearly understood. For
instance, the recent numerical studies in [12]-[13] assumes afixed depth of moving soil, in which the movement of soil issimultaneously mobilized along the pile. However, in apractical scenario, soil movement may be graduallymobilized to a deep layer. The effect of progressive soil
movement on the response of piles has not been completely
investigated, particularly, when it is coupled with axial load.
With an experimental apparatus developed, extensive tests
have been undertaken for piles and pile groups in sand toinvestigate the responses of piles under combined verticalload and lateral soil movement. Reference [5] presents simple
solutions for piles in moving sand from the results of 14
typical 1g model pile tests. References [7]-[8] further verified
the solutions and findings by investigating the pile responses
due to effective soil movement and impact of a uniform andtriangular soil movement profile.
This paper investigates the group effects of piles due to lateral

2. APPARATUS AND TEST PROCEDURES
For brevity, detailed information regarding the apparatus, test
preparation, data acquisition and measurement system, anddata process program are omitted. Only the relevant parts arebriefly introduced herein.
2.1 Shear Box and Loading System
A schematic cross section of the shear box and the loading
system is shown in Fig. 1. The internal dimensions of theshear box are 1 m by 1 m, and 0.8 m in height. The upper partof the shear box is made of a series of 25 mm thick square
laminar steel frames. The frames, which are allowed to slide,contain the “moving layer of soil” of thickness L m . The lower
section of the shear box comprises a 400 mm high fixed
timber box and the desired number of frames to achieve a“stable layer of soil” of thickness L s (≥ 400 mm). A loading
block is used to apply lateral force on the laminar frames,
which is made into arc, rectangular, and triangular shape to
impose varying soil movement profiles. In this paper, an arc
loading block was used. The rate of movement of the upper
shear box (thus the soil) is controlled by a hydraulic pump(lateral jack), and a flow control valve. A vertical jack is used
to install the pile into the shear box.
2.2 Model Piles
Fig. 2 shows a schematic diagram of the model pipe pileunder testing. The aluminum pile has a length of 1200mm,
outer diameter of 32mm and wall thickness of 1.5mm. The
pile was instrumented with 10 pairs of strain gauges at an
97interval of 100mm. Prior to the testing, the strain gauges were
calibrated by exerting a transverse load in the middle of thepile that was clamped at both ends. Given various magnitudesof the load, measured strains were compared with those
calculated for each gauge, so that a calibration factor was
obtained, which in turn allows a measured strain to beconverted to an actual strain. To protect from damage, thegauges were covered with 1mm epoxy and wrapped by tapes.

2.4 Test Procedures
The model tests were conducted using an arc loading block asshown in Fig. 2. The instrumented pile was installed at a
distance, S b of 500mm from the loading side (see Fig. 1(b)) toan embedded depth, L of 700 mm in the shear box. The
predetermined final sliding depth, L m and stable layer depth,
L s was maintained as 200mm and 500mm. After pile
installation, a pile cap fabricated from solid aluminum wasused to secure the piles in the pile cap. Weights were put on
the pile cap to apply axial load. The strain gauges wereconnected to a data acquisition system and measurement
system established. During testing, the frame movement, wf
was measured from a reference board using a ruler moving
with the top frame (see Fig. 1(b)) and up to 120mm. Each test
is denoted by a combination of letters and numbers, e.g.
AS32-0 or AG32-3d-294, where ‘AS’ or ‘AG’ signifies
Single pile or pile Group test using an Arc loading block; ‘32’
indicates 32 mm in diameter, ‘3d’ denotes the pile center tocenter spacing, S v =3d, ‘0’ or ‘294’ represents an axial load of
294N or 30kg per pile.

2.3 Sand Properties
The sand used in this study was oven dried medium grained
quartz, sand with a uniformity coefficient C u of 2.92 and acoefficient of curvature of C c of 1.15. The sand wasdischarged into the shear box through a rainer hanging overthe box to achieve a reasonably uniform density of the sandwithin the shear box. The falling height of sand was chosen as
600 mm, which gave a uniform relative density of about 89%,
and a unit weight of 16.27kN/m3. Angle of internal friction
was 38° as evaluated from direct shear tests.
Height adjustablebridgeVerticalcolumns

Vertical jack
500

Strain gauges

wf
Lm=200

Frames

A
Lm

A

25

LVDT

LVDT

100mm

Model pileLs : stable layer depth
unit: mmFixed timber box

100 50(a) Elevation view
Reference board for measuring wf

Verticalcolumn3 TEST RESULTS

500 mmSoil movement

Figure 2. Schematic of a tested pile

Pile B
Pile ASb

500 mm

500 mmPile cap

The pile deflection profiles were derived from double
numerical integration of the bending moment profiles, using
the measured pile deflection and rotation at ground surface as
input boundary conditions. The shear force profiles werededuced by single numerical differentiation of the bending
moment profiles. It is noted that both piles A and B (Fig. 1(b)were instrumented with strain gauges. The pile responses
obtained from the two piles are similar, therefore only thedata collected from pile A were used. These profiles wereplotted at every 10mm frame movement, wf for all the tests.
Due to space limitation, only the profiles in tests AS32-0 and

500 mmLateraljack

Ruler moving
with top frame

Sv

Horizontal
beam

Loading
block

Ls=500Ls

Lm : predetermined final
sliding depth

100

d =32
400mm

Lateraljack(b) Initial plan view (A-A)
Figure 1. Schematic of shear box
98AG32-3d-0 were presented. The development of maximum
bending moment, M max , shear force, T max and pile deflection
at ground line, y 0 are plotted against the frame movement, wf
for all the tests. Table 1 summaries these values at wf=120mm.

impact on the development pattern of M max , T max and y 0
with w f .
0(a) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

Depth (mm)200

3.1 Test AS32-0
Test AS32-0 was conducted without axial load. Similar to thetests using a triangular loading block described in [5], the arcloading block induces an increasing soil movement bothhorizontally and vertically with the progressive mobilizationof the frames. Fig. 3 shows the pile response profiles for aframe movement, wf up to 120 mm. The bending moment
profile is of parabolic shape when w f ≥90mm. The bending
moment increases significantly at a frame movement wf≥80mm and has not reached the ultimate value even at w f= 120mm. The measured maximum bending moment occurs
at about the same depth of 350 mm, which is about half of thepile’s embedded length. At wf =120mm, two local largest
shear forces of -137.2N and 122.4N was deduced from thebending moment profile. The pile rotated around the pile tip
with a pile deflection of 4.2mm.

300
400
500
600
700

0

10

20

30

40

50

Bending moment, M (kNmm)

0(b) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

Depth (mm)

200
300
400
500
600
700
-200

-100

0

100

200

Shear force, T (N)

3.2 Test AG32-3d-0
Test AG32-3d-0 was conducted with the pile arranged in arow (S v =3d) perpendicular to the direction of soil movement
and pile head capped as shown in Fig. 1. Presented in Fig. 4
are the five response profiles. A very similar trend to thoseobserved in the single pile test TS32-0 was noted. At wf =120mm, the maximum bending moment, M max , is 23.5
kNmm, occurring at a depth of 350 mm. Again, two largest
shear forces, T max of -96.2N and 88.8N occur in the sliding
and stable layers, respectively. The pile deflected by rotation,with a magnitude of about 3.6mm at the ground level.

0(c) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

Depth (mm)

200
300
400
500
600
700

-1

0

1

Soil reaction, p (N/mm)

03.3 Response of Mmax, Tmax and y0 versus wf

100(d) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

200
Depth (mm)The development of the maximum bending moment, M max
shear force, T max and pile deflection at ground level, y 0 for thefive tests are plotted in Figs. 5(a), 5(b) and 5(c) against theframe movement, wf . These figures demonstrate:• The initial frame movement wi of 40mm caused negligible
or little M max , T max and y 0 for the pile group tests and single
pile test using the arc loading block. At wf =40~80mm, thethree critical responses for the single pile are generally lessthan those of the piles in a group, but overall the pile
responses are still very small. Afterwards, the single pile
responses surpass those of the pile groups, showing greaterrigidity of the pile groups with the pile head capped atlarger soil movement.
• The development pattern of M max , T max and y 0 with w f forthe pile groups and single pile are similar, revealing thatthe pile cap and pile-soil-pile interaction have insignificant
impact on the performance of the two piles in a row.
• Imposing the axial load of 588 N on the pile cap, i.e. 294 N
per pile, and the pile spacing of 3d and 5d have little

300
400
500
600
700

0.00

0.01

0.02

0.03

Rotation, S (radian)
0100(e) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

Depth (mm)

200
300
400
500
600
700
-6

-4

-2

0

2

Deflection, y (mm)

99

4

6

8

Figure 3. Response of pile during AS32-00(a) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

Depth (mm)

200
300
400
500
600
700

0

10

20

30

Bending moment, M (kNmm)

0(b) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

Depth (mm)

200
300
400
500
600
700
-120

-80

-40

0

40

80

120

Shear force, T (N)

3.4 Maximum Pile Response Profiles
The largest bending moment, shear force and pile deflection
profiles with depth for the five tests at wf =120mm are plottedin Figs. 6(a), 6(b) and 6(c). These figures show that
• The bending moment profiles of the pile group tests andsingle pile test are similar amongst themselves andanalogous to parabolic. The maximum bending moment
occurs at a depth of 350mm below the ground surface,which is 0.5 times of the pile embedded length.
• The shear force profiles are also of similar shape. Twolocal largest shear forces of approximately equalmagnitude but of opposite signs are deduced from thebending moment profiles.
• The bending moment and shear force profiles of the single
pile encompass those of the pile group tests.
• The axial load of 588N and pile spacing of 3d and 5d havelittle impact on bending moment and shear force profiles,
except that very small bending moment of 1.67kNmm and
4.0kNmm was measured at the ground level in tests
AG32-3d-294 and AG32-5d-294.
• Pile deflects mainly by rotation about the pile tip or a depth
near the pile tip.

0200
Depth (mm)4 DISCUSSIONS(c) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

100

300
400
500
600
700
-1

0

1

Soil reaction, p (N/mm)

0100(d) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

Depth (mm)

200
300
400
500
600
700

0.000

0.004

0.008

0.012

0.016

Rotation, S (radian)

0100(e) wf (mm)
10
20
30
40
50
60
70
80
90
100
110
120

Depth (mm)

200
300
400
500
600
700
-4

-2

0

2

4

4.1 Mmax versus Tmax
The deduced T max in Fig. 5(b) are plotted in Fig. 7 against thecorresponding measured M max in Fig. 5(a) for the five tests.
This figure shows a remarkably good linear relationshipbetween T max and M max under any w f and independent of thepile head restraint conditions and pile spacing.
Reference [5] presents equivalent elastic solutions for piles
subjected to moving sand, giving
M max =T max L/m(1)
where L(=0.7m) is the embedded length of the pile and m is a
non-dimensional constant. The value of m is estimated as 2.8
with a variation of ±5% for the best curve fitting between T max
and M max for the 32mm diameter pile tests using a triangularloading block. The current test results corroborate this linear
relationship as shown in Fig. 7. This equation may be used in
the estimation of the lateral thrust (shear force), which is
required in the design of reinforcing piles to increase slope
stability in [15].
4.2 Group effect
In order to assess the group effect and investigate the pile-soil
interaction behavior of piles subjected to lateral soil
movement, the critical pile responses of a pile within a group,
such as the maximum bending moment, pile head deflection
and limiting soil pressure are compared with those of a single
pile. Investigation on group effect on the behavior of piles
subjected to lateral soil movement has been carried out both
numerically and experimentally on piles with various pilehead fixity conditions in sand and weathered soil in[2],[4],[16]. Reference [4] demonstrates that the group effectDeflection, y (mm)Figure 4. Response of pile during AG32-3d-0100quantified in terms of measured maximum bending moment
may be more reliable and consistent.

those obtained from the previous studies on capped-piles insand. The current test results indicate that for the capped-piles

Figure 6. Response of pile during AS32-0

Figure 5. Response of pile during AS32-0

Maximum shear force, Tmax (N)150

Measured data
AG32-3d-0
AG32-3d-294
AG32-5d-0
AG32-5d-294
AS32-0100

Mmax=TmaxL/2.8
50

0

0

10

20

30

40

Maximum bending moment, Mmax (kNmm)

Figure 7. Measured Mmax versus Tmax

Figure 8. Variation of Fm with pile spacing Sv/dIn this study, the group effect was evaluated by a group factor,
F m based on the measured M max .
Fm =M g max
M s max(2)

in which M gmax is the maximum bending moment for a pile ina group at a frame movement, M smax is the maximum bending
moment from the single pile at the same frame movement.
Table 1 summarize the group factors, F m at wf =120mm andin a row, the group factor F m for the group with a spacing of
3d is less than those of 5d. The axial load of 294 N per pile has
virtually very limited impact on the group factors. The groupfactors are plot in Fig. 8 against the normalized pile pacing. It
is found that the group factors are less than unit in theinvestigated pile spacing of (2.5~7)d and decrease as the pile
spacing decreases. Nevertheless, the group factors from thecurrent results are 17% and 30% less than those presented in101[2][16]on average. This may be attributed to the pile positionsin the shear box discussed below.
Table 1 Summary of test results at wf =120mm
Test
AS32-0
AS32-3d-0
AS32-5d-0
AS32-3d-294
AS32-5d-294

M max(kNmm)
36.6
23.5
24.3
22.9
24.5

T max
(N)
122.4
88.8
90.7
85.3
100.0

6 REFERENCES

y0
(mm)
3.6
5.0
3.6
4.4
4.8

Table 2 Summary of group factor Fm
Spacing
Current (without load)
Current (with load)Reference [16]Reference [4]

2.5
0.82
0.72

3
0.64
0.63

5
0.66
0.67
0.94
0.78

6. The group factors, F m from current experimental
investigation are 17% and 30% less than those obtained in[2],[16].

7
0.96
0.84

The experimental results from three series of nine tests on a
25mm and 32mm pile using a triangular and rectangularloading block [5]-[7] show that the M max decreases withincreasing distance S b between the pile location and theloading side where free soil movement is generated. The
experimental investigation on two piles in a row subjected tosoil movement induced by a triangular loading block in [6]further reveals that the group factor F m decreases withincreasing S b for both free and capped pile head fixity
conditions.
5 CONCLUSIONS
Laboratory model tests were conducted to investigate theresponses of piles due to lateral soil movement. The results ofa single pile test and four group tests on two piles in a row
with pile head capped were presented. These tests werecarried out using an arc loading block, which induces
progressive soil movement. The group effect was assessed byusing group factors. The key findings from the studies are
summarized as follows:
1. The bending moment, shear force and deflection profiles
for a pile in the two piles in a row group tests are very
similar to the shape of those in a single pile test, including
the position of the maximum bending moment, but theirmagnitudes are generally smaller, showing group effect.
2. The development pattern of M max , T max and y 0 with wf forthe pile groups and single pile are similar, revealing that thepile cap and pile-soil-pile interaction have insignificant
impact on the performance of the two piles in a row.
3. The axial load of 588N on the pile cap, i.e. 294N per pile,
has little impact on the responses of the piles in a row.
4. A linearly relationship exists between the M max and the
T max , which can be described by Eq. (1) with m=2.8.
5. The group factors, F m are less than unit in the investigatedpile spacing of (2.5~7)d and decrease as the pile spacing
decreases.

[1] Steward DP, Jewell RJ, Randolph, MF, “Design of piled
bridge abutment on soft clay for loading from lateral soil
movement,” Geotechnique, vol. 44, no. 2, pp. 277–296.
[2] Chen LT, “The effects of lateral soil movements on pilefoundations,”Ph D Thesis, University of Sydney,
Australia, 1994
[3] Poulos HG, Chen LT, Hull TS, “Model tests on single
piles subjected to lateral soil movement”, Soils Found.,vol.35, no.4, Dec. 1995, pp.85-92.
[4] Chen LT, Poulos HG, Hull TS, “Model tests on pilegroups subjected to lateral soil movement”, Soils Found.,vol.37, no.1, Mar. 1997, pp.1-12.
[5] Guo WD, Qin HY, “Thrust and bending moment for
rigid piles subjected to moving soil,” Can. Geotech. J.,
vol.47, no.2, Feb 2010, pp.180-196.
[6] Qin HY, “Response of pile foundations due to lateralforce and soil movements,”Ph D thesis, Griffith
University, Australia, 2010.
[7] Qin HY, Guo WD, “Pile response due to lateral soil
movement of uniform and triangular profiles,” in
GeoFlorida2010, 2010, pp. 1515-1522.
[8] Qin HY, Guo WD, “Pile response subjected to effective
soil movement,”in Proc. 7th Int. Conf. on PhysicalModelling in Geomechanics (ICPMG) 2010, pp.
823-828.
[9] White DJ, Thompson MJ, Suleiman MT,Schaefer VR,“Behavior of slender piles subjected to free-field lateralsoil movement”, J.Geotech. Geoenviron. Eng., vol. 134,
no. 4, Apr. 2008, pp.428-436.
[10] Frank R, Pouget, P, “Experimental pile subjected to long
duration thrusts owing to a moving slope,”
Geotechnique, vol. 58, no. 8, Aug. 2008, pp. 645–658.
[11] Smethurst JA, Powrie W, “Monitoring and analysis of
the bending behavior of discrete piles used to stabilize arailway embankment,”Geotechnique, vol. 57, no. 8, Aug.
2007, pp. 663–677.
[12] Kourkoulis R, Gelagoti F, Anastasopoulos I, Gazetas G,“Slope stabilizing piles and pile-groups:parametric study
and design insights,” J.Geotech. Geoenviron. Eng., vol.
137, no. 7, July 2011, pp.663-677.
[13] Kanagasbai
S,
Smethurst
JA,
PowrieW,“Three-dimensional numerical modeling of discrete piles
used to stabilize landslides”, Can. Geotech. J., vol.48, no.
11, July 2011, pp.1393-1411.
[14] Guo WD, “p u based solutions for slope stabilizing piles,”Int. J. Geomech., 2011, to be published.
[15] Poulos HG, “Design of reinforcing piles to increase slope
stability”, Can. Geotech. J., vol.32, no.5, May 1995,
pp.808-818.
[16] Jeong S, Kim B, Won J, Lee J, “Uncoupled analysis ofstabilizing piles in weathered slopes,” Comput. Geotech.
vol. 30, Aug. 2003, pp. 671-682.

102

[17] Viggiani C, “Ultimate lateral load on piles used tostabilize landslide,” in Proc. 10th Int. Conf. on SMFE,
1981, 555-560.

103

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Collapse-Settlement Calculations for Embankments Based ona Collapsible Soil Subgrade
Moshe Livneh, Technion-Israel Institute of Technology, Haifa, Israel
Noam A. Livneh, Noam Livneh Engineering Ltd., Haifa, Israel
ABSTRACTCollapsible soil deposits are found extensively in many arid and semi-arid regions of the world. These soils areconventionally considered problematic, since they exhibit a high potential for collapse under loading. Upon loading, allsoils settle, the amount of settlement varying from soil to soil and dependent on load-induced stresses; but such
settlement will eventually cease after a certain period of time. However, under certain conditions, subsequent wetting
may cause additional settlement, known as collapse. Evaluation of the collapse potential is, thus, required for settlement
calculations relating to the base of embankments placed on a collapsible soil subgrade. The evaluation conducted here
leads to the conclusion that collapse potential is dependent upon the following parameters: (a) liquid limit, (b) plasticity
index, (c) in-situ dry density, (d) in-situ moisture content, and (e) applied pressure at wetting in the single-odometer
test. Knowledge of these parameters, together with that of the water-penetration depth into the subgrade stratum and theeffect of partial saturation on a reduction of the full collapse potential, is shown to enable a calculation of the amount ofcollapse-settlement.
Keywords: Collapse, Embankment, Settlement, Silty-characteristics
equilibrium deformation at each level of pressure. It is theninundated at a prescribed applied pressure, and thedeformation is measured. The deformation induced by theaddition of water, divided by the initial height of thespecimen, expressed in percentage terms, defines the collapse
potential, which this paper also terms vertical collapse.
Now, a prediction of the settlement of the base of anembankment as a result of added water, and thus as a result ofsubgrade collapse, necessitates knowledge of thecollapse-pressure characteristics (curves) of the silty strata
under consideration. These curves, which are usuallyobtained from laboratory tests on undisturbed and remolded
silty samples, are in general terms dependent on thefollowing parameters: (a) liquid limit, (b) in-situ moisture
content, (c) in-situ dry density, and (d) coefficient of
uniformity.
In light of all these introductory notes, the objectives of thispaper are as follows:• Developing the collapse model for a given data oflaboratory collapse tests conducted on silty
undisturbed and remolded specimens.
• Formulating the effect of partial saturation oncollapse potential and evaluating the in-situ
saturation distribution from wetting.
• Outlining the suggested collapse-calculation
procedures for any given site conditions.
Finally, through an Excel spreadsheet, the paper offers apractical example of how the suggested collapse calculation
can be applied with any given set of data.

1. INTODUCTION
Collapsible soil deposits are found extensively in many arid
and semi-arid regions of the world [1]. These soils areconventionally considered problematic, since they exhibit ahigh potential for collapse under loading. Upon loading, allsoils settle, but the amount of settlement varies from soil tosoil and is dependent on load-induced stresses. Althoughsuch settlement will eventually cease after a certain period oftime, subsequent wetting under certain conditions may cause
additional settlement, known as collapse. An evaluation of asoil’s collapse potential should, thus, be included in all
settlement calculations. In this paper, the base ofembankments placed on a collapsible soil subgrade provides
the objective of these calculations.
According to [2], soils subject to collapse have ahoneycombed structure of bulky shaped particles or grains
held in place by a bonding material or by capillary-tension
forces. Common bonding agents include soluble compounds,
such as calcareous or ferrous cementation, that can be
weakened or partly dissolved by water, especially acidic
water. Removal of the supporting material or force occurs
when water is added, enabling the soil grains to slide or shear
and move into voids.
Two methods currently exist to determine the amount ofcollapse in the laboratory, the single-odometer test and the
double-odometer test. This paper refers only to thesingle-odometer test, known also as ASTM D 5333 collapse
testing. In this test, an undisturbed or remolded specimen is
first driven into the odometer ring and then subjected toincreasing vertical load. The specimen is permitted to attain

2. VERTICAL COLLAPSE MODEL
As mentioned previously, predicting the settlement of an103

embankment-base surface as a result of subgrade collapse
necessitates knowledge of the collapse-pressure characteristics(curves of the collapse model) of the silty strata underconsideration, and these are usually obtained from laboratory
tests on undisturbed silty samples. This prediction model
denotes a statistical relationship between the vertical collapse(Cp), in percentage, that is due to applied vertical pressure
(Pp)
in
kPa,
and
the
followingparameterscharacterizing the silty specimens under examination:
moisture content (W) in percentage, in-situ dry density (D) in
kN/m3 and dry density at liquid limit (D LL ) in kN/m3 as
defined by Eq. (2). For 73 local specimens extracted from asilty subgrade situated in the southern part of the country, forwhich the plasticity data are given in Figure 1, thebest formulation of this model is as follows:
Cp=28.5354-27.0305×(D/D LL )0.9825++0.0001196×[log(Pp)]11.3741/(W/PL)1.4908(1)

D LL =9.807×100/(100/G+LL)

(2)

R

R

P

R

P

It is interesting to compare all the statistics given above withthose in Table 1. This table contains the statistical criterion
for goodness of fit (SE/SY) and validity of correlation (R2) as
taken from [6] for five states of correlation: excellent, good,
fair, poor and very poor. The comparison leads to theconclusion that the Excel Solver analysis of the 65 specimens
given in Fig. 1 can be categorized as leading to a fair state ofcorrelation for Eq. (1) together with Eq. (2).
P

Table 1: Statistical criterion for goodness-of-fit statistic(SE/SY) and validity-of-correlation statistic (R2), after [6]Criteria for
R2
SE/SY
Correlation
Excellent
≥0.90
≤0.35
Good
0.70-0.89
0.36-0.55
Fair
0.40-0.69
0.56-0.75
Poor
0.20-0.39
0.76-0.90
Very Poor
≤0.19
≥0.91
P

P

P

R

P

P

P

P

P

In addition, Fig. 3 depicts Eq. (1) graphically for the silty
material. For comparison purposes, this figure also includes
the collapse results obtained from the laboratory testing of the
73 specimens of Fig. 1.
5T

In Eq. (2), G denotes the solid specific gravity of the silty
specimen, and LL denotes its liquid limit in percentage. Hereit should be noted that Eq. (1) has been derived with the aid of
the Excel Solver command.

5T

25

18
20

Measured vertical pressure [%]y = 0.7819x - 12.573
R² = 0.9124

16

Plasticity index [%]

14
12
10
8
6

A-Liney = 1.0092x - 0.0445
R² = 0.5350
15

10

5

0
-10

-5

0

4

5

10

15

20

-5
Undisturbed specimens

2

Remolded specimens

Undisturbed specimens
Remolded specimens

0
20

22

24

26

28

30

32

34

36

-10

Predicted vertical collapse [%]Figure 2: Measured vertical collapse versus predicted verticalcollapse from Eq. (1) for the experimental data shown in Fig. 1

38

Liquid limit [%]Figure 1: Plasticity data from 73 silty specimens (65undisturbed and 8 remolded)

30

As shown in Fig. 2, the R2 (coefficient of determination)
value obtained for the collapse model of Eq. (1) is 0.535. In
addition, the accompanying SE (standard error) value
obtained is 4.206%, and the SE/SY (ratio of standard error tostandard deviation of the measured vertical collapse values)
is 0.707. In other words, 68 percent of all predictions from
Eq. (1) may be expected to be accurate within ±4.206%, and95 percent to be accurate within ±8.412%.
Here, it should be mentioned that Eq. (1) was formulated toyield the highest R2 value from among all possibilities that (a)include the independent variables of the soil's plasticity and(b) yield true physical behavior as clearly described in [3](i.e., an increase in Cp with increasing Pp, and a decrease in
Cp with increasing W, D and LL). The D/D LL variable of Eq.
(1) exactly follows this finding as given in [4] or indirectly as
given by a similar finding in [5].
P

Undisturbed Pp=200 kPa
Undisturbed Pp=100 kPa
Remolded Pp=200 kPa
Combined Pp=200 kPa
Combined Pp=100 kPa

P

W/PL=0.25

25

Vertical collapse [%]

P

20

0.50
0.25 0.75
0.50

15

0.75
10

P

R

5

0
0.0

0.5

1.0

1.5

2.0

Ratio of in-situ dry density to dry density at liquid limit

Figure 3: Vertical collapse versus density ratio (D/D LL ) for
vertical pressures of 100 kPa and 200 kPa as derived from Eq.
(1) for the experimental data shown in Fig. 1

R

R

104

R

Finally, in order to illustrate that Eq. (1) yields the highest R2value obtained from among various regression formulations,the following equation may be offered for comparison
reasons:
Cp=1,009.1849-944.1516×(W/PL)0.004543-95.0980×(D/D LL )0.8239+16.6917×[log(Pp)]0.8239
P

R

R

P

predicted Cp values might result from the fact that most of thespecimens for Eq. (4) are of the undisturbed type and all ofthem for Eq. (5) are of the remolded type. For lower values ofCu, the percentage values change to a lower range of ratio
values.

P

P

P

P

(3)

30
Undisturbed Pp=200 kpa
Undisturbed Pp=100 kPa
Remolded Pp=200 kPa
Combined Pp=200 kPa
Combined Pp=100 kPa

2For this equation, the R value obtained is lower, 0.510. In
addition, the accompanying SE obtained is higher, 4.318%,
as is the SE/SY obtained, 0.726.
P

25

Verical collapse [%]P

3. ALTERNATIVE COLLAPSE MODEL
In connection with Eq. (1) and Eq. (3), another multiple linear
regression equation can be derived for the data of Fig. 1. Thisequation, although it yields a somewhat higher value for
R2, 0.595 (see Fig. 4), and a somewhat lower value for SE,
4.049%, as well as a somewhat lower value for SE/SY, 0.650,
contains only two independent variables:
P

P

P

P

10%
5%
10%

15%
20%

10

15%
20%

0
10

(4)

12

14

16

18

20

In-situ dry density [kN/m3]

Figure 5: Vertical collapse versus dry density ratio for verticalpressures of 100 kPa and 200 kPa as derived from Eq. (4) forthe experimental data shown in the Fig. 1

The outputs of Eq. (4) are shown in Fig. 4 and Fig. 5. The
latter figure indicates that its predicted Cp values are
somewhat lower than those given in Fig. 3. It should be notedthat Eq. (4) follows the general pattern of Eq. (5), reproduced
from [7]. For this latter equation, the R2 value obtained for its
experimental data (138 remolded specimens) was muchhigher than the preceding two R2 values, 0.94. Furthermore,
the SE value was much lower than the preceding two values,
1.94%. Finally the formulation of this equation is as follows:
P

W=5%
15

5

P

Cp=26.8201-0.5065×W-2.6995×D+10.9653×log(Pp)20

Following Eq. (4), the results obtained for additional multiple
linear regressions conducted with (a) the X 1 , X 2 and X 3
independent variables are shown in Table 2 and (b) the X 4
independent variable, which as always is equal to log(Pp).
R

P

R

R

R

R

R

R

R

P

Cp=48.496+0.102×Cu-0.457×W-3.533×D++6.447×log(Pp)

Table 2: Summary of statistic-values of four multiple linear
regression analyses conducted on vertical collapse testingdata for 73 specimens
X1
X2
X3
SE [%]
SY/SE
R2
W
D
LL
3.806
0.640
0.614
W/PL: D/D LL LL
3.809
0.640
0.613
(1)
)
W
D
--3.8690.650
0.595
W/PL D/D LL--4.327
0.727
0.493
(1)
This row represents the outputs for Eq. (4).

(5)

R

25

R

R

P

R

Measured vertical pressure [%]

20

P

P

R

y = x + 0.0000
R² = 0.5947

15

P

Table 2 indicates that the best statistical results are associatedwith the equation in the first row below the heading row; i.e.,
Eq. (4) modified by adding the LL variable). However, thisequation cannot be regarded as applicable, as it leads toincreased collapse values with increasing LL values. The
same applies to the equation in the second row below theheading row. In other words, these findings lead to theconclusion that the plasticity inputs in the regression analysis
should be eliminated.
This conclusion is compatible with [8], who suggests
eliminating the use of LL or Pl in these regression analyses
and, instead, including the gradation characteristics, such as
the percentage difference between gravel and fines, and thepercentage difference between sand and fines. Thissuggestion is of similar magnitude to that given in [7, 9], who
include sand percentage, silt content and clay percentage intheir regression analyses. All these findings lead to theconclusion that the regressions of Table 2 with the LL
variable should be eliminated; as a result, Eq. (4) is seen to be

10

5

0
-10

-5

0

5

10

15

20

-5

Undisturbed specimens
Remolded specimens-10

Predicted vertical collapse [%]Figure 4: Measured vertical collapse versus predicted verticalcollapse from Eq. (4) for the experimental data shown in Fig. 1
Compared with Eq. (4), Eq. (5) contains an additional
independent variable, Cu, which denotes the coefficient of
uniformity of the prescribed silt. For the case of Cu= 80,
which represents the average Cu value for the 73 specimensof Fig. 1, Eq. (4) yields higher Cp values than those of Eq. (5)by about a ratio of 2.1-3.9 for the W range of 5-15%, D=13
kN/'m3 and Pp=200 kPa. For the Pp=100 kPa case, this rangeof ratio values increases to 2.2-13.6. These differences in the
P

P

P

105

the most favorable of the four equations given in the table.

The calculation of collapse for field conditions should
obviously take into consideration the distribution of thedegree of saturation with depth after wetting. According to[11], the depth of a change in moisture content occurring inthe silty subgrade of an unpaved shoulder in a semi-arid zone
resulting from a cumulative total of 300 mm rainfall per year
extends down to 2,000 mm only. Fig. 8 depicts this moisture
distribution as reported in [11]. An analysis of this
distribution with the aid of Eq. (8) and Eq. (9) leads to theconclusion that a good correspondence exists between theamount of effective rainfall that has penetrated into the silty
subgrade and the characteristics of the moisture distribution
shown in the figure.

4. FINAL COLLAPSE MODEL
Finally, a determination needs to be made of the regression
equation--the collapse model--that is more preferable toemploy, Eq. (1) or Eq. (4). To do so, it is advisable first to
check the differences in the predicted vertical collapse values
between these two equations. Figure 6 depicts the predictedvertical collapse value of Eq. (4) for the given 65 specimens
compared to Eq. (1). As this figure indicates, the results ofthese two equations are, on the average, similar. Yet, as Eq.
(4) does not contain any independent variable to represent theplasticity of the specimens examined, it seems that Eq. (1) ispreferable one, but only if the findings of [4] and [5] onplasticity influence are to be adopted.

1.0
0.9
0.8

y = 0.9874x
R² = 0.8143

15

0.7

Collapse reduction

Predicted vertical pressure from Eq. (4) [%]

20

10

5

0.6
0.5
0.4
0.3
0.2

0
-10

-5

0

5

10

15

20

0.1

-5

0.0

Figure 6: Predicted vertical collapse from Eq. (4) versus
predicted vertical collapse from Eq. (1) for the data of the 73
specimens of Fig. 1

P

Ro=-6.95×(ΔSo) +7.20×(ΔSo) -0.20×(ΔSo)-0.004151(6)

ΔS O =(S F -S I )/(100-S I )

(7)

R

R

R

R

P

R

R

R

P

0.5

0.6

0.7

0.8

0.9

1.0

R

5

10

15

20

25

Point D

Depth below subgrade surface [m]Point D

0.4

Point C
Point C

Point C
Point D

0.8

1.2

1.6
Before the rainfall season

After 152 mm rainfall
After 300 mm raifall
2.0

Figure 8: Moisture-content versus depth as measured in asilty subgrade from an unpaved shoulder of a flexible
pavement, after [11]The same pattern of moisture-content or degree of saturation
distribution is reported in [8]. This reference deals with afull-scale test in which artificial wetting (by ponding the site)
is applied to cause settlement collapse in a silty stratum
possessing a single concrete footing on its surface. For thistest, the degrees of saturation monitored before and after
wetting are given in Fig. 9, in which the degree of saturation
distribution, as suggested by this paper for design purposes, isalso included.

In Eq. (6), Ro denotes the collapse reduction; i.e., the ratio ofvertical collapse resulting from a partially saturated state to
the full vertical collapse resulting from a fully saturated state.
ΔSo denotes the ratio increase in the degree of saturationfrom wetting as defined by Eq. (7). In the latter equation, S I
denotes the initial degree of saturation (prior to wetting) inpercentage and S F denotes the final degree of saturation (after
wetting) in percentage.
R

0.4

Moisture content [%]0

Partial wetting of the specimens in the single-odometer test
results in only a proportion of the total collapse value that iscompatible with the case of total specimen inundation. The
study conducted in [10] indicates that full collapse essentiallyoccurs for degrees of saturation of 65%-70% and above.
However, for degrees of saturation of 50%, only about 85%of the full collapse occurs.
Fig. 7, reproduced from Fig. 14 in [10], enables formulationof the aforementioned collapse reduction resulting frompartial saturation. The equations obtained are the following:
P

0.3

0.0

5. EFECT OF PARTIAL WETTING2

0.2

Figure 7: Collapse reduction (Ro) versus increased ratio in the
degree of saturation (ΔSo)-10

Predicted vertical collapse from Eq. (1) [%]3

0.1

Increased ratio in the degree of saturation

Undisturbed specimens
Remolded specimens

R

Equation (5)Figure 14 of [10]

0.0

R

R

106

In addition, the data in [10] also includes in-situ
moisture-content distribution before and after artificial
wetting. Figs 9-11 in this reference indicate that the general
pattern of moisture-content distribution or the degree ofsaturation distribution in Fig. 8 or Fig. 9 also exists in thesefigures.

wetting is:ΔS OZ =ΔS OT ×(Z R -Z)/(Z R -Z C )
R

20

40

60

80

R

R

R

R

R

R

(13)

R

R

100

0
Data taken from Table 2 of [8]500
Test Hole 1

Depth from surface [mm]R

The development of Equations (12) and (13) enables acomputation of settlement resulting from rain water
penetration into the silty stratum for any given in-situ data.
This computation is performed by dividing the silty stratum
into horizontal strips down to a depth of Z F ) (i.e., usually,again, 3,000 mm). The final computed settlement (δ) for ahomogeneous silty stratum is derived from the following
equation:

Degree of saturation [%]0RR

δ=Σ(Roi×Cpi×di)/100

1,000

Where: di denotes the thickness of the horizontal strip of thesilty stratum under consideration (usually, a constant value of
500 mm, thus 6 strips in all for the case of Z F =3,000mm); Roidenotes the collapse reduction for the middle depth of the ith
strip; and Cpi denotes the vertical collapse in percentage,again for the middle depth of the ith strip, for which thevertical pressure (Ppi) is given by the following equation:

Test Hole3

Test Hole 4

(14)

1,500

R

2,000

R

P

Test Hole 2

2,500

Before wetting
After wetting
Suggested distribution

P

3,000

R

R

R

R

6. SETTLEMENT CALCULATIONS
The first step in calculating the settlement of an embankmentbase constructed on silty subgrade is to evaluate S F and ΔSofrom the rainfall quantity together with the final moisture
distribution as described in Fig. 9. This calculation is
performed with the aid of the following equations:(8)

W FT =2× Z F ×W FM /(Z C +Z F )

(9)

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

S I =100×W I /(100×9.807/D I -100/G)
R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

(15)

R

R

R

R

R

P

P

P

Base-Settlement Calculations
Spreadsheet for Embankements & Homogeneous Strata
29-27.03046499 0.982485

011.3741

1

Design Input for the Silty Srtartum

R

Design Input for the Embankement

8

Liguid Limit [%]30

Height of Embankement [mm]9

Plasticity Limit [%]20

Thickness of Pavement [mm]9In-Situ Moisture Content [%]

10

10.0

3

In-Situ Dry Density [kN/m ]14.7

10 Thickness of Compaction [mm]
11

0.00
####

600

Collapse Model Coefficients

12,000 0.0

Wet Density of fill material [kN/m 3]
3

Wet Density of Pavement [kN/m ]

600

2.4

18.7

1.8

22.7

0.9

Rainfall & Moisture Data per Season 0.2
150

Effective Rainfall per Year [mm]

##

R

R

R

R

R

R

P

P

P

Calculated Intercept Value

1.64

12

Calculated Slope Value

12
3

3.3

5.0

32.9

Depth of Active Zone [mm]3,000 0.0

0.0003Depth of Uniform Moisutre [mm]1,000 0.0

49.4

Cal' Increase in Top Saturation [%]

0.25

##
##

Calculations

R

Depth from
Surface ofSubgrade to
Middle Depth of
Horizontal Strip

VerticalPressure on
Middle of
Horizontal
Strip

Full
VerticalCollapse

Calculated
Ratio
Increase in
Saturation.

Final
Partiall
VerticalCollapseAmount ofPredicted
Stllement

[mm]

[mm]

kN/m 3

[%]

[ /.]

[%]

##

300

242.9

0.0

0.2

0.0

0

##

##

850

251.8

8.8

0.2

2.4

12

##

R

R

PR

R

11

R

R

R

R

P

In these equations, W I denotes the initial moisture content inpercentage, W FM denotes the average final moisture content
along the wetted zone in percentage, W FT denotes the final
moisture content at the top of the subgrade in percentage, Z F
denotes the depth of active (wetting) zone in mm (i.e., usually
3,000 mm; see Fig. 9), Z C denotes the depth of the constantwetting region measured from the surface of the subgrade in
mm (i.e., usually 1,000 mm; again see Fig. 9), H W denotes
the effective rainfall quantity that penetrates into thesubgrade in mm, D I denotes the dry density of the silt before
wetting in kN/m3, D F denotes the dry density of the silt after
wetting in kN/m3 , G denotes the solid specific gravity of thesilt, S I denotes the initial degree of saturation (prior towetting) in percentage, and S FT denotes the final degree ofsaturation (after wetting) at the top of the subgrade inpercentage. Note that these equations refer to a homogeneous
silty stratum in regard to in-situ water content and dry density.
Similar equations, however, can be developed for aheterogeneous silty stratum.
Now, for the subgrade depth range of 0 to Z C mm, theincreased ratio in the degree of saturation that is due to
R

P

P

(11)

R

R

R

R

(10)

R

S FT =100×W FT /(100×9.807/D F -100/G)
R

P

R

W FM =100×H W /(Z F ×D I /9.807)

R

Where: H E denotes the height of the given embankment in
mm, D E denotes the wet in-situ density of the embankment in
kN/m3, H P denotes the thickness of the given pavement in
mm, D P denotes the wet in-situ density of the pavement in
kN/m3, Zi denotes the depth to the middle thickness of the ith
strip, and Ppi denotes the vertical pressure acting at the
middle thickness of the ith strip in kPa.
R

RP

Ppi=H E ×D E +H P ×D P +Zi×D I ×(1+W I /100)

Figure 9: Degree of saturation versus depth as measured in
the collapse field test of [12]]

P

R

##

1,350

259.8

9.2

0.2

1.8

9

##

##

1,850

267.9

9.7

0.1

0.9

4

##

##

2,350

276.0

10.2

0.1

0.2

1

##

##

2,850

284.1

10.8

0.0

0.0

0

##

##

3,350

292.2

11.3

0.0

0.0

0

##

##

3,850

300.3

11.8

0.0

0.0

0

##

##

4,350

308.4

12.4

0.0

0.0

0

##

##

4,850
-6.95

316.4
12.9
7.2
### -0.0042

0.0

0.0

0

##

R

R

R

R

R

R

107

Sum of Predicted Settlements [mm]27

Figure 10: An Excel spreadsheet demonstrating a settlement
computation for the data given in the figure
The best way to demonstrate the settlement computation is byPpresenting a specific illustrative example. This was done withthe aid of the Excel spreadsheet shown in Fig. 10 (located in
the previous page) for the data given in the uncolored cells of
the figure and for the collapse model formulated by Eq. (1).
To emphasize, the computations given in Fig. 10 assume
these distribution inputs for the rainfall and moisture:
H W = 150 mm, Z C = 1,000 mm and Z F = 3,000 mm. The final
settlement result is 27 mm, which is an acceptable value forwithstanding the collapse event. For the same Fig. 10 data,
the use of Eq. (4) instead of Eq. (1) leads to a lower total
settlement value: 25 mm. Thus, the use of Eq. (1) seems to bepreferable in order to enhance the factor of safety.

spreadsheet for a given practical example.
8. ACKNOWLEDGEMENT
The paper is based on engineering studies conducted for
Olizki Earthworks, Roads & Development Ltd. underassignment by Israel Railways Ltd., and thanks are therefore
due them.
9. REFERENCES[1] Wiseman, G., and Lavie, G., "Arid Zone Subsurface
Exploration and Loading Tests," Proc. of the 7th AsianRegional Conf. on Soil Mechanics and Foundation
Engineering, Vol. 1, 1983, Haifa, pp. 98-104.
[2] U.S. Army Corps of Engineers, "Application with
Unstable Foundation Soils," Chapter 5 in Engineeringand Design Settlement Analysis, EM 1110-1-1904,
1990, pp. 5-1 to 5-18.
[3] Lawton, E.C., Fragaszy, R.J., and Hardcastle, J.H.,"Collapse of Compacted Clayey Sand," Journal of
Geotechnical Engineering, Vol. 115, No. 9, 1989, pp.
1,252-1,267.
[4] Zur, A., and Wiseman, G., "A Study of Collapse
Phenomena of an Undisturbed Loess," Proc. of the 7th
Int. Conf. on Soil Mechanics and Foundation
Engineering, Vol. 2.2, 1973, Moscow, Russia, pp.
265-269.
[5] Austerlitz, G., Ishai, I., and Komornik, A., "Prediction ofCollapse Potential in Israeli Loess Subgrade," Proc. of the
7th Asian Regional Conf. on Soil Mechanics andFoundation Engineering, Vol. 1, Haifa, 1983, pp.
106-111.
[6] Dongré, R., Leslie, M., D’Angelo, J., Paugh, C., andGudimettla, J. "Field Evaluation of Witczak and Hirsch
Models for Predicting Dynamic Modulus of Hot-Mix
Asphalt," Journal of the Association of Asphalt Paving
Technologists, Vol. 74, 2005, pp. 381-442.
[7] Basma, A.A., and Tuncer, E.R., "Evaluation and Control
of Collapsing Soils," Journal of GeotechnicalEngineering, Vol. 118, No. 10, 1992, pp. 1,491-1,504.
[8] Junk, C.H., and Elton, D.J., "Predicting Collapse
Potential with Neural Networks," Transportation
Research Record 1582, 1997, pp. 22-28.
[9] Basma, A.A., and Kallas, N., "Modeling Soil Collapse
by Artificial Neural Networks," Geotechnical andGeological Engineering, 22, 2004, pp. 427-438.
[10] El-Ehwany, M., and Houston, S.L., "Settlement andMoisture Movement in Collapsible Soils," Journal of
Geotechnical Engineering, Vol. 116, No. 10, 1990, pp.
1521-1534.
[11] Livneh, M., and Shklarsky, E., "Saturation Criteria in
Pavement Design for Semi-Arid Zones," Proc. of the Int.
Conf. on Moisture Equilibria and Moisture Changes inSoils beneath Covered Areas, 1965, pp. 237-247.
[12] Houston, S.L., Houston, W.N., and Spadola, D.J.,"Prediction of Field Collapse of Soils Due to Wetting,"
Journal of Geotechnical Engineering, Vol. 114, No. 1,
988, pp. 40-58.

7. SUMMARY AND CONCLUSIONSThis paper dealt with the issue of calculating collapse
settlement, specifically that of the base of an embankment
constructed on collapsible silty subgrade. An essential input
for this type of calculation, in addition to such routine
characteristics as liquid limit, plasticity limit, in-situ moisture
content and in-situ dry density, is an evaluation of thecollapsing characteristics of the silty stratum; that is,, thevertical collapse percentage.
It was shown that for a given set of experimental results on
undisturbed and remolded silty specimens, Eq. (1) constitutes
the best available collapse model of all models analyzed,
utilizing both the Excel solver command technique and themultiple linear regression technique. This equation includes
plasticity inputs, whereas such inputs were found to beinadequate in the multiple linear regression analysis.
The accuracy of all collapse models examined was found to
be only fair: i.e., more than a poor state but less than a goodstate. This is a frequent finding, however, in similar analyses
offered in the technical analysis literature [10].
In addition, it should be recalled that the collapse model isbased on experimental results obtained from ASTM D 5333
collapse testing (also known as single-odometer testing).
Therefore, this model is compatible only with a final wetting
state of full saturation. However, it was shown that only a
final state of partial saturation is reached. The influence of
this partial final site condition state on vertical collapse
values is very significant as shown in Fig. 7 taken from [10].
Thus, the formulation of the curve given in this figure yields
an important input for the settlement calculations discussed.
In order to utilize the reduced vertical collapse resulting frompartial saturation, the depth of the wetting zone and thedegree of saturation distribution along this zone should be
evaluated. This was done in this paper by analyzing threedifferent site tests reported in the technical literature [10, 11
and 12]. Moreover, it was found that the amount of effective
rainfall that penetrates into the subgrade governs thecharacteristics of this distribution.
The final conclusion is that all the above findings can serve asa proper tool for calculating the collapse settlement of thebase of an embankment constructed on a collapsible silty
subgrade. This is illustrated in this paper by utilizing an Excel
108

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Study on the influence of soil mechanical properties of earth pressure cell
installation trench on stress measurement and local arching in earthfill dam tocalibrate total stress registered by earth pressure cell
D. Elmi1, A. A. Mirghasemi2
School of Civil Engineering, College of Engineering, University of Tehran, Tehran, Iran
ABSTRACTAn understanding of the total stress distribution in an earthfill dam plays a key role in monitoring and guaranteeing dam
safety. The registered total stress by earth pressure cell is not enough satisfactory and many factors such as installation
method complicate its performance. To avoid damage of pressure cell, it is recommended to backfill the installation
trench with fine grained and light compacted soil. This nonconformity in soil mechanical properties of the trench resultsin local arching between the core and trench. The presence of installation trench changes total stress distribution, andthe registered pressure in presence of trench becomes less than the pressure in its absence. In design calculation, thisregistered stress must be calibrated. This paper studies the changes of local arching by increasing the elevation of dam
construction to find methods for stress calibration.
Keywords: Arching, Earthfill dam, Earth pressure cell, Stress measurement
1 INTRODUCTION
Geotechnical monitoring of earthfill dam during
construction and operation is critical to control static anddynamic stability of dam. For this purpose different
instruments are used in a dam. Earth pressure cells are
installed within earthfill dams to monitor stress in different
directions. The accuracy of their measurements is important
for anticipating destructive phenomena like hydraulicfracture in an earthfill dam. Earth pressure cells are difficult
to monitor accurate total stress due to various reasons suchas installation method. Dunnicliff and Green (1988) haveexpressed their concerns on factors affecting earth pressure
cell readings such as cell dimension, stiffness ratio, aspectratio and placement effects [1].
A report by the International Society for Rock Mechanics(ISRM, 1981c) includes detailed recommendations for
installation of embedment earth pressure cells [2]. The
report recommends excavating a trench with a compacted
and level base. Then, cells should be installed in small
pockets at the base of excavation and protruding stones
should be removed since earth pressure cells are prone todamage due to heavy loads. Although it is ideal that the soil
within the trench have a density similar to that of the core,
it should be backfilled using stone-free soil and compacted
by hand-operated equipment.
Although the above procedure prevents damage of cells but
many studies have been done to increase the accuracy ofmeasurements. Matsuu et al. (2008), quantitatively
evaluated measurement errors caused by embedding
method and the trench geometry for earth pressure cells to
be installed in the central core-type rock-fill dam [3].
Ahangari and Noorzad (2010) carried out laboratory tests,
using a casing material for the cell to improve the accuracyof the results. Their study shows that when the casingmaterial is softer than surrounding soil, the pressure cellunderestimates the stress [4]. Shahbazian and Fakharian(2007) indicated that measurement of total stress is not only
affected by the inclusion made by the pressure cell, but the
trench geometry and installation procedure also contributes
to nonconformities in the medium. They concluded that themain dominating parameter in installation is over/undercompaction of a backfill material [5].
The accuracy of stress measurement is evaluated by archingratio, which is the proportion of registered stress totheoretical normal stress:

ArchingRatio =

σv
γ .z

(1)

Where σ v is total vertical stress registered by pressure cell, γ
is soil unit weight and z is earthfill height above thepressure cell.
In an ideal measurement, when the stiffness of medium isequal with the surrounding, arching ratio is near to unity.
Conversely, smaller arching ratio means the stress istransmitted from flexible zone to stiffer one.
Three types of arching are probable in an earthfill dam:First, transversal arching which happens between different
zones (shell, core, filter,...) of an embankment dam. Second,
longitudinal arching which is caused by steep slope ofvalley, usually V shape. This occurs between body and dam
abutments. Third, local arching that is possible when the fillaround pressure cells has different material compared tosurrounding medium.
Presence of pressure cell installation trench with different
material affects natural stress distribution. Total stresstransmits from the trench to the core and pressure cellregisters less stress at this location. This inaccuracy of fieldreadings should not be considered in designing program.
This paper, studies on changing of local arching during
construction steps to calibrate total stress data of pressure
cell.

109

2 NUMERICAL MODEL
Bulk mod ofIn this study the installation trench is modeled within an fluid (GPa)earth fill dam in ABAQUS which uses finite element
Friction
method to analyze the problem. Fig. 1 shows a 2D model ofthe non-homogeneous earthfill dam in which dam height, angle (deg)core width and shell slopes are assumed to be 128m, 66m
Dilationand 1:2.5 respectively. The trench is situated 4 meters angle (deg)above the bed. To simulate a real model with reliable
Cohesionresults, dam sides are assumed to be 600m and bed depth is(kg/cm2)
two times of dam height.
P

2

2

-

-

2

20

20

35

39

39.4

0

0

3

5

9

0.3

0.3

0

0

0.85

0.35

0.35

0.45

0.3

0.4

P

Void ratio

A basic concept in ABAQUS is the division of the problem
history into steps. As depicted in Fig.3 dam is constructedin 46 layers within 2 years. Layers are generally 4 meters at
except at upper elevations and near to the crest. In
construction steps the load changes linearly from one
magnitude to another.

Fig. 1 The view of dam-foundation model
Soil is assumed to be 100% saturated with Mohr Columb
behavior. Table 1 presents soil mechanical properties of
dam in different zones. To mesh the model, 4-node bilinear
plane stress element (CPS4) is used. Fig. 2 shows the gridmesh zone for the dam.

Fig. 3 dam construction time
The choice of initial time step is important in consolidation
analysis in ABAQUS. Due to the coupling of spatial andtemporal scales, it follows that no useful information isprovided by solutions generated with time steps smaller
than the mesh and material-dependent characteristic time.
This issue is discussed by Vermeer and Verruijt (1981) [6],who propose the criterion:

Fig. 2 Quadratic mesh for different regions

Table I. Material properties of earthfill dam

Density(kg/cm3)
Young’s
Modulus(MPa)
Poisson’s
Ratio
P

TrenchCore

Transition
zone

shell

1740

1740

1950

2050

35

35

70

102

0.35

0.35

0.27

0.25

Kx
=1e-8
Ky
=1e-9

K x =1e-8
K y =1e-9

1e-5

1e-6

P

Permeability
(m/s)

R

R

R

R

R

R

R

R

∆t ≥
bed

γ w .∆h 2
6 E.k

(1)

Where
is the distance between nodes of the finite
element mesh near the boundary condition change, E is theelastic modulus of the soil skeleton, k is the soil
is the specific weight of water.
900 permeability, andTo verify that the geostatic stress field is in equilibriumwith the applied loads and boundary conditions on the
0.25
model, initial stress condition of bed is considered. In thiscase, the loads and initial stresses equilibrate and produce
1e-10 zero deformations. Zero-valued boundary conditions are
imposed on displacements to constrain the movement of theselected degrees of freedom of bed to zero. To specify thezero pore pressure at the surface of each layer during
2100

110

construction, pore pressure boundary conditions are
imposed in each step.
3 ANALYSES RESULTS
To clarify the influence of mechanical properties on local
arching, the proportion of trench to core was set to 1/2, 1/5,
1/10, 1/20, 1/35, 1/50 and 1/100.
In non-homogeneous earth fill dam, transversal arching
happens between various zones due to nonconformity ofmaterial. By completing the construction steps coresettlement increases and more stress transmits from flexible
zone (core) to stiffer one (shell). During initial steps ofconstruction up to 10-15 m backfilling, the influence oftransversal arching is negligible and just local arching
happens between the core and the trench due to different
material. By increasing dam elevation, registered total
stress by pressure cell will be affected by local andtransversal arching. Fig. 4 represents total stress
distribution along a horizontal path in dam section atpressure cell elevation in different steps. It is clear that at
the end of construction stress is significantly transmitted
from core to the shell, while it is more slightly at stages of 9
and 13m of backfilling.

arching. In other step S1 subtracted from S2 to evaluate
intensity of local arch action (ΔS L ).
The proportion of measured stress to theoretical stress(ΔS L /Theoretical Stress) presented for the center of trench,in different stiffness, during the construction procedure to
specify arching ratio changing. Fig. 5 shows local archingratio (ΔS L /Theoretical Stress) for the trench in 7 modes of
stiffness.

Fig. 5 Local arching ratio at the trench center, during
construction
As it is shown in Fig. 5, when the elastic modulus of thetrench was half of the surrounding medium, the transmitted
stress to the core is about 5% of the normal pressure above
the cell (γ.h) during backfilling. Analyses show that local
arching remains approximately constant during
construction procedure. In this case, it is possible tosubtract local arch action from field readings of pressure
cell to calibrate data and reach the natural field of stress indam body.
3 CONCLUSION

Fig. 4 Total stress distributions along a horizontal path inconstruction steps
Evaluation of stress and transversal arching is critical toguarantee safety of embankment dam; however, presenceof trench changes the natural stress field. To reach thenatural field of stress in presence of trench, calibration
should be done on pressure cell data for designing
procedure.
A study was taken on local arching ratio during
construction steps to express total stress calibration. For
this, the stiffness of trench material set equal to stiffness ofthe core. Thus, registered stress by pressure cell (S1) was
affected by transversal arching. In other analyses, theproportion of elastic modulus of trench to core material set
to 1/2, 1/5, 1/10, 1/20, 1/35, 1/50 and 1/100 and stress
distribution plotted for trench center with respect toconstruction level (S2). In this case, measured stress at
center of the trench was affected by local and transversal

Earth pressure measurements are carried out in order todetermine stress redistribution occurring within the dam.
The reliability of these measurements is significantly
reduced, however, by many factors such as presence ofinstallation trench with material different from the core.
Therefore, local arching changes the natural stress field andtransmits total stress from trench to stiffer zone (core).
This paper studied on local arching trend by increasing theconstruction elevation. The study concluded that the stresswhich was transmitted to the core, due to stiffness
nonconformity, remained approximately stable during
construction steps. Moreover, subtraction of pressure cell
field readings from analytical stress in absence of trenchmaterial is a constant percent of normal stress (γ.h).
Therefore, local arching can be divided from field readings
to reach to natural stress in absence of the trench.
4 REFERENCES[1] Dunnicliff, J. and Green, G. E. ”Geotechnicalinstrumentation for monitoring field performance,”111

John Wiley and Sons Inc. New York. 1988,
ISBN-0-471-09614-8, pp 165-177[2]ISRM (1981c),”Suggested Methods for Pressure
Monitoring
Using
Hydraulic
Cells,”
Rock
Characterization Testing and Monitoring, ISRM
Suggested Methods, Pergamon Press, Oxford, 1981, pp.
201-211[3] Matsuu, M. Takashi, A. Uchita, Y. and Ohta, H.
“Improvement of earth pressure measuring method inrock-fill dams,” International commission on large
dams 76th annual meeting, 2-6 June 2008,Sofia,
Bulgaria.
[4] Ahangari, K. and Noorzad, A. “Use of casing and its
effect on pressure cells,” Mining Science andTechnology, 20(2010) 0384-0390.
[5]Shahbazian, H. and Fakharian, K. “Numericalinvestigation of trench geometry and soil property
effects on the accuracy of embedded pressure cellmeasurements,” OttawaGeo2007: The Diamond Jubilee
Conference, 2007, Ottawa.
[6]Vermeer, P. A. and Verruijt, A. “An Accuracy
Condition for Consolidation by Finite Elements,”
International Journal for Numerical and Analytical
Methods in Geomechanics, Vol. 5, 1981, pp. 1–14.

112

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Studying the effects of considering dilation angle over the performance ofgeotechnical elements using numerical methods
Salman hanifi1, Amin naeimabadi2 , Mohammad reza atrchian 3
Islamic Azad University Of Zanjan, Iran
ABSTRACT
Applying geo textile and nail elements adjacent to structures like: shallow foundations, dams, tunnels, slopes, retaining
walls, etc is a very important and inevitable case in most geotechnical projects. In other hand, installing these elements
truly and precisely, needs the distinction of geological and geotechnical condition in project site. Among the countlessmechanical soil parameters, dilation angle is one the most effective and also controversial parameters, so that any
highly over/under estimation of this parameter could result to irrecoverable structural and economical consequences. In
this paper the effects of considering the dilation angle over performance of above elements in soil are studied, using thefinite element code FLAC.2D. Also further to considering a great variation of quantities for dilation angle and gaining areasonable trend, comprehensive results are acquired. It is observed that variation of dilation angle between φ/3 up to
φ/2(ϕ is the soil internal friction angle) has a great effect on geo textile elements, so that increasing of dilation angle
leads to a more stable condition in geotechnical structures containing elements like geo textiles or nails.
Keywords: Dilation angle, Geo textile, Nailing, FLAC 2.D

1. INTRODUCTION
If the adjacent soil of a geotechnical structure does not haveproper condition and durability/ resistance, various ways suchas inserting geo synthetic layers, soil stabilization, and nailingcan be used. In addition, among the geo synthetics, geo
textiles are the best and most useful ones. These plates are
similar to fabric layers which instead of cotton, wool and silk,
synthetic materials are used in their textures. At least 80
specific uses are defined for geo textiles, but their basic
application can be summarized in five cases: Separation,
reinforcing, filtration, drainage and sealing membrane[1]. By
entering geo synthetics to geotechnic engineering, a lot of
research was done by researchers such as Seki (1986), Korner(1994), Basarest (1998), sheen and Das (1998). They studied
different kinds of geotechnical structure buried in reinforced
soil by geo textile material. Furthermore in recent years otherstudies about geo synthetic behavior available in shallow
were done by shookla and yeen (2006), Chang andkasekant(2007) and Sawav(2007).
Three decades ago, nailing was introduced as a technique fortrench stability and protection of excavation, wall nailingmeans reinforcing existing soil in the form of in situ andinstalling adjacent steel bars in a sloppy area or dug inground. In this way the bars usually are put in to holes createdin soil walls and are controlled in their own place by cement
grout. This process has passive performance and applies its
effect by soil-bar interaction. The issue of soil interaction
with reinforced elements has been in the focus of many
researchers. Clearly, true understanding of this problem
necessitates information about soil mechanic parameters andtheir effect on interaction between soil and elements. One of

the most important parameters is dilation angle. As we know,
in general, five parameters are needed in Mohr-Coulomb
behavioral model which are as follows: Young modules, void
ratio, friction angle, cohesion and dilation Angle. Dilation
angle has a unit of degree. For heavily-over consolidated
clay, the dilation angle is counted zero. On the other hand theamount of dilation in sands depends on compaction andfriction angle. for quartz sand is approximately φ-30 and for
φ<30 it's amount can be considered zero. Additionallynegative amount of ψ can only be applied for very loose
sands. By using finite element program Afena, Vafaiean[2]calculated bearing capacity coefficients of shallow
foundations in wide spectrum of parameters including
dilation angle. Fahimifar and Javaheri[3], examined shear
behavior and dilation of two sliding surfaces on each other bymodeling direct shear test. Rabi'ee & Shahkarami[4] byexamining the effect of various parameters on bearing
capacity of foundation on the slope, found that by increasing
dilation angle from zero to φ, the amount of bearing capacity
increased such that they recorded the effect of dilation angle
on less slope, in a clearer and better way. Fredman andBurd(1997)[5] examined the changes of bearing capacity
coefficients (N γ ) for dilation angle between zero to φ in thesmooth and rough foundations and came to the conclusionthat in smooth foundations and for internal frication angle
under 35º, changing the dilation angle, significant difference
is not observed in N γ amount, but for amount of friction angle
plus 35º and change of dilation angle between 0.0 to φsignificant changes is observed in bearing capacity
coefficient N γ . Shahbarkhordar & et.al)[6] examined thesensitivity of bearing capacity coefficients N γ toward changesof dilation angle and by using finite element method. By
calculating bearing capacity coefficients, Griffiths [7]observed that in the case of dilation angle equal zero,113

Significant instability is created in calculation process(trends). Owlaapour & Fakhradini [8] worked on the effectsof various parameters such as dilation angle and emphasized
numerical analyses the effect of clay layer on bearing
capacity of sandy soil by using finite element program
PLAXIS. Amir Hossein & Ali shaff'ie[9], also calculated
bearing capacity coefficient of a shallow strip foundation for
wide variety of changes in dilation angle. In the present paper,modeling and examining the performance of geo technicalelements, geo textile and nailing operation by using finite
difference program (FLAC.2D) was emphasized. The
performance of these two geo technical elements is calculated
for different amounts of soil dilation angle and by following areasonable trend, general results are achieved.
2. Introducing FLAC.2D software
Flac is a two-dimensional finite difference program that is
exclusively used for geotechnical calculations. by using thissoftware, we can model the behavior and performance of any
construction (structure) on soil , stone or any building
materials which may undergo a kind of plastic flow during theyield limit. in a general view, materials are defined in FLAC
by help (means) of elements or Zones, And by forming a
network, the geology shape of the object or structure to bemodeled, is formulated. Both elements in this software aredefined based on a linear or non-linear stress-strain rule andreact against applied boundary conditions. Another feature of
FLAC is yielding and flowing of materials in parallel tochanges in network form in the case of large-strain mode.
Accurate calculation of Langrage along with techniques
based on definition of behavioral zones and availability ofwide variety of materials in this software, makes themodeling of failure and flowing more exact (accurate). On theother hand, since no matrix will be formed in FLAC, it ispossible to do two-dimensional calculations without the needfor extra memory. Also formulating (existing limitation in thegiven time to run the problem or definition of needed
damping) have been lifted in this virtual space especially byautomatic scaling inertia capability and defining automatic
damping (this problem has no effect on failure situation).
Although FLAC has exclusively been designed to solve geo
technical and mining engineering problems, it has also great
capability in analyzing complex mechanical problems. Suchthat by means of various defined behavioral models in it wecan model and simulate many of geologic non-linear andirreversible behaviors.
3. statement of the problem
3.1. soil (ground) space
The target soil in the problem site, was mono layer sand andhad the following specifications:

Table 1: constant parameters used in the modeling trend of
soil space.
Resistancecohesion(C) kPa
0.0

Internal
frictionangle(φ)
35

Bulk
modulus
kPa
2×108

Shear
modulus
kPa
1×108

Unit
weightKN/m3
10

We know that, introducing the parameter of dilation angle in
the position of internal friction angle which is greater than orequals 30 is meaningless and can be ignored. Therefore inorder to have a realistic modeling the degree of internal
friction angle of target sandy soil in this study is 30 to 40 andto be more exact it is 35. It must be noted that the target
failure criterion in this modeling was Mohr-Coulomb criteria.
3.2. Geo textile Element
In order to have a much better and effective geo textile
element, two parameters should be taken into account. First,
the question of flexibility of texture of geo textile and second,the question of the interaction of inserted geo textile with its
own upper and lower soil surface. Performing this task infinite difference program FLAC.2D, is possible throughconnecting beam element to its upper and lower networklines. In this way by putting moment of inertia of modeled
beam to zero in the FLAC.2D consider its performance asequal to geo textile elements. furthermore, one of the major
problem of soil profiles reinforced with geo textile element, isthe possibility of sliding on both sides of installed geo textile
plates and consequently is the happening of pull out andeventually paying the way for large strain.

Fig 1: schematic drawing of investigated geo textile model
According to fig(1) the used numerical model is a rectangle
network in the dimensions of 5(m)*7(m) [5 meter width 2 &7 meter depth). The boundary conditions on this model are
regarded in such a way That the bottom and the upper part ofthe model is supported in both directions, and its left and rightends are free in both directions. The reason for leaving free
support conditions in both ends, as to facilitate the modelingof sliding phenomenon. In this way by inserting geo textile
element in the center of target profile, It is possible toobserve the effect of pull out happening on it Directly. In
order to apply loading in the case of strain control, the load
has been introduced as application of velocity vectors. The
amount of these velocity vectors in vertical loading equals

1140.5×10-4 and the case of loading it is regarded 1×10-4 m foreach step of loading. The amount of velocity vector in bothstates, is considered in such a way that, it prevents an initialshock to system and to neutralize and modify the initial andtemporary effect of the loading.
3. 3. Nailing Operation

Element network for geo textile modeling is based on fig 3.
Having done the modeling, the problem for dilation angles
equals Zero, φ-30(5 degree), φ/3 (about 12 degree), φ/2(about 18 degree) and φ (equals 35 degree), is solved. It
should be noted that the assumption of equality between
amounts of dilation and internal friction means that soil
follows an associated flow rule which is not a correct
assumption. But in the present article, to compare with theresult of other targeted amounts, for dilation angle andexamining the effect of increasing the amount of ψ, we havetaken this assumption. Based on Table 2 and Fig 4, we can seethat by increasing dilation angle, the amount of bearable axial
force on part of geo textile element is increasing, in otherwords its bearing capacity is rising.
Table 2: The amount of axial force applied on geo textile on
the moment of failure
35103547

Fig 2: schematic drawing of nailing model.

ϕ/2≈1884954

ϕ/3≈12
59237

5.0
48657

0.0
42358

ºΨ
Fa(N)

Modeling of nailing operation in ELAC.2D has been done bycable element According to figure (2), the applied model is arectangle network with the dimensions of 11m×11m(11 meter
width * 11 meter depth). In this model border lines are
considered in such a way that model bottom is supported inboth directions but its left and right end(s) has reactions only
in horizontal direction or(X axis). In this situation three layersof nailing are predicted for protection of a vertical deposit. In
this case the behavior and reaction of nail elements areanalyzed against the force of deposit weight. In order to makereal the modeling of nail elements, we can show the effects ofgrout injection around the nails by considering a friction
resistance among the nail and sand soil. This resistance force
is regarded as a friction angle between nail and soil and its
amount, is determined 20 according to reference[10]. On theother hand, by regarding support conditions by maintaining
more consistency in nail performance, the effect of shotcrete
performance has been taken into account.

In table (2) the amount ψ dilation angle is based on degree,
and F a is the amount of axial Force on geo textile element interms of Newton at the moment of failure. As mentioned
before, by increasing dilation angle, the bearing capacity ofgeo textile element is also increased. According to figures 4
and 5, The greatest amount of the increase has been recordedin ψ equals to 12 to 18 degrees.

4. Modeling and analyzing the results

Table 3: The resultant increasing amounts in axial bearing
capacity according to increasing the dilation angle

RR

Fig 4: Changes of amounts of axial force according to dilation
angle.

4.1 geo textile element

18-35
22

12-18
43

5-12
22

0.0-5
15

Ψº
Δ(%)

In table 3, Δ is increasing in amount of axial bearing capacityof geo textile element in term of percentage. As it can be seen,
the greatest amount of this increase is in ψ within 12 to 18
degree which record in ψ/3 to ψ/2. Additionally It was seenthat by increasing dilation angle, the depth of shear failure
mechanism is also increased. In fig 5 and 6 we can easily see
the progress of shear failure mechanism, particularly underinserted geo textile layers.
Fig 3. Element network of geo textile problem in FLAC.2D

115

angles and internal friction implies that soil acts as associatedflow rule, which is not a true assumption, but in this paper it isaccounted to compare with results of other targeted amount ofdilation angle and assessing the effect of increasing theamount of ψ.
According to table 4 and fig 8, we can see that by increasing
dilation angle, the amount of bearable axial force is increasedon part of each of the nails.
Table 4: Resultant increasing amounts of axial bearing
capacity of nails regarding to increasing the dilation angle
Fig 5: Shear failure areas (zones) under geo textile regarding
ψ=12oBased on fig 5 and 6, It can be seen that the depth of shear
failure mechanism in dilation angle equals 12 degrees, is lessthan the situation in which dilation angle is 18 degrees. In
other words in ψ=18º, bearing threshold of system is
evaluated much greater than ψ=12 degrees , such that byforming greater failure zones, the system undergoes
instability and failure.

35

18

12

5

0.0

21687

18553

14924

11761

9439

27379

22491

18325

15587

12325

29953

23284

20136

17693

16554

Ψº
Cable
1
Cable
2
Cable
3

Fa
(N)

Fig 8: Variation in amounts of axial force of nails according
to dilation angle
According to Table 4, the amount of increase in bearing
capacity, has directly relationship with increasing dilation
angle and the depth of buried nail element, such that nail orcable number 3 with the greatest amount of depth regarding
dilation angle of 35 degrees experiences (bears) the greatest
amount of axial force. [The nail are put in order of 1, 2 and3
from top to bottom]. According to Table 5, by increasing
buried depth, the increase in bearing axial capacity of nails inψ between φ/2 to φ is very conspicuous. put simply, the more
the amount of surcharge soil is increased, the greater amountof dilation angle (with in φ/2 to φ) will have more effect onenhancing axial bearing capacity of nails. It is noted againthat in Table 5, Δ is increasing in axial bearing capacity ineach of nails in terms of percentage.

Fig 6: Shear failure zones under geo textile regarding ψ=18º
4.2. Nailing operation

Table 5: Resultant increasing amount in axial bearing
capacity regarding to increasing the dilation angle

Fig 7: Element network for nailing problem in FLAC.2D
The applied element network for modeling of nailing
operation is given in fig 7. After modeling, just like what wasdone before, the problem is solved regarding dilation angles
equal to Zero, φ -30 (5 degrees), φ/3 (about 12 degrees), φ/2(about 18 degrees) and φ (35 degrees). It must be noted once
again that the assumption of equality the amounts of dilation

116

18-35
17
22
28

12-18
24
23
16

5-12
27
18
14

0.0-5
25
25
7

Ψº
Cable 1
Cable 2
Cable3Δ(%

However for lower amount of dilation angle, increasing ofburied depth, not only did not cause the increase of bearing
capacity of nails (relative to upper nail) but also bring about
decreasing bearing capacity of nail elements in comparison
with its upper elements.

International Congress on Civil Engineering, May 11-13
2009, Shiraz University, Shiraz, Iran.
10. FLAC.2D manual, Itasca Consulting Group, Inc.
Minneapolis, Minnesota, USA.
11. Feyzabadi, Mohiyadin, shahnazari, Habib, Numerical
Modeling of shallow foundation next to a slope 4thinternational geotechnical engineering and soil mechanic, 2-3
November 2010 Tehran, Iran, Code, and number of article:
678 (MANFEY).
P

5.CONCLUSION
In this paper the effect of dilation angle on the performance orfunction of geo technical elements like geo textiles andnailing operation was examined. Modeling of these two cases
was done by finite difference code FLAC.2D and in thecondition of Mohr-Coulomb criteria. Generally it was seenthat by increasing dilation angle, the axial bearing capacity ofgeo textile and nails is also enhanced. In other words, byincreasing dilation angle, the threshold of system failure andthe width (extension) of shear failure zones are increased;
meaning that by accounting greater amounts for dilation
angle, the necessary condition for instability in system isforming and joining more failure zones. Furthermore, inmodeling geo textile element, the greatest amount in bearing
capacity occurs in dilation angle within φ/2 to φ/3. However
the problem of increasing amount of bearing capacity inmodeling the operation of nailing does not follow a specificrule or order. We come to conclusion that due to greater
amount of dilation angle (within φ/2 to φ) as the amount ofsurcharge soil increases, the axial bearing capacity is also
intensified.
6. REFERENCES
1. Designing with geosynthetics. Robert.M.Koerner .
2. VaFaaiean Mahmoud: Determining bearing capacity
coefficient of soil by means of finite element and based on
Mohr-coloumb elastic model, esteghlat publication, year:22,
No 1, 2003.
3. Fahimifar, Ahmad, Amin Javaheri, Amir Reza, Modeling
direct shear experiment for study the behavior of fissuresaround tunnel. transportation research Quarterly 8th year
second Vol 2011.
4. Rabi'ee, Meysam, Amir Shaakarami, A, Examining theeffect of Various parameters on bearing capacity offoundation over a slope.
5. Frydman, S., and Burd, H., "Numerical studies ofbearing-capacity factor N γ ", Journal of geotechnical and geo
environmental engineering., vol 123, No 1, pp. 20-28, 1997.
6. shahbarkhordar, S.H, Roshanzamir, M.A, Hashem
Alhoseyni.H, Calculating the coefficient of N γ in bearing
capacity by using finite element method, The first national
civil engineering congress April 2004. San'ati Sharif
university, article code 83-1193.
7. Griffiths, D. V .,"Computation of bearing capacity factors
using finite elements", Geotechnique. 32. No. 3, pp.195-202,
1982.
8. Owlapoor, M. Fakhradini, M. Numerical Analysis of theeffect of clay layer on bearing capacity of sand soil 4th
international geotechnical engineering&soil mechanic, 2-3
November 2010 Tehran, Iran. Article Code: 432 (BATOLA)
9. Shafi'ee, Amir Hossein, Shafi'ee, Ali, "Computation of
bearing capacity factors using finite element method", 8th
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R

R

P

P

P

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Comparing the excavating effect of single and twin tunnels on the earth surface
settlement by numerical methods
Amin Naeimabadi , Salman Hanifi , Mohammad reza Atrchian
Department of Geotechnic Engineering, Faculty of civil engineering, Islamic Azad University, Zanjan Branch, Zanjan, Iran

ABSTRACT
Nowadays due to increasing rate of urbanization, population compaction and construction in Iran, the issue of
construction of underground tunnels become the center of attentions, due to it's capability to obviate the problems
regarding transportation, water transmission and subsurface structures. In this way, studying about the right placement,
procedure of excavating tunnels and it's possible effects on superstructures specially an ancient buildings are very
important and vital. In this research, the settlement of a single tunnel excavated in the depths of Isfahan city (Cultural
capital of Islam world) in Iran has was studied, using the finite element program PLAXIS, then the results of thisnumerical analyze was compared with the data obtained from field instrumentations, Finally these results weregeneralized to twin tunnels. In the next stage, after studying the effect of spacing between twin tunnels over these
settlements, those quantities, which are greater than the allowable settlement and their situation were recognized. It is
obvious that this problem will highlight the importance of Following a good and reliable design process.
Keywords: Settlement, Twin Tunnels, Numerical Method, PLAXIS

1. INTRODUCTION
The approach to growing use of tunnels to facilitatetransportation in busy and crowded parts of cities has brought
about deflection in soil mass and eventually in groundsurface. Among the features of urban tunnels, their lesssurcharge depth, loose ground of towns, large span of theseareas and the buildings on the stations are important, All ofthem make the control of shallow settlement and maintaining
stability in the underground spaces very difficult. Settlements
due to tunnel excavation, can cause serious damages to theneighboring structures.
In this study, we examined single tunnels by numericalmethod and twin tunnels by computer modeling and accurate
instrumentations. The researches on comparing the effect of
single tunnel and twin tunnels on settlement of ground surface
has been limited. In Iran, after the growing trend of subway
tunnel excavation, in urban regions, irreparable damage ofground surface settlement on building and ancientmonuments particularly in Isfahan and Siyo-Se-poul, arenoticeable [Fig 1] .

Fig 1: Damage to Siyo-Se-Poul, a worldly recordedmonument due to settlement of ground surface.

Perhaps Terzaghi's report[1] is among the first reports thatshow the effect of soil type in transferring settlement toground surface. Regarding coarse-grained soil, he believedthat all ground displacement in tunnel site owing to dilation
gradually disappear to ground surface. However in clay layer,
this effect is less clear mostly due to lack of volume change
and soil layer which are influenced by cohesion.
Peck(1969) presented a comprehensive report containing
general collection of executive operation of tunnel in soft
ground. In his detailed examination, he divided settle able
ground into four categories. Then for each of soils mentioned,
measurements for tunnels situated in these soils[12]. Heshowed that the cure of ground surface settlement to tunnel
excavation has normal distribution and presented a relation
for the profile of settlement as follows:x2(1)
S ( x ) = S max × exp( )2iThis equation has no theoretical foundation and chosenbecause of its similarity to experimental form of settlement.
In i and S max were determinant and a lot of research was done
to determine them amounts and many suggestions were madeto calculate them. Most of the studies are to modify determine
i. Another important parameter is S max [Fig2]. To achieve a
relationship to calculate the amount of this parameter, aparameter called ground loss is defined ,which is in fact
equals to contraction of tunnel span. In clay soils, thecondition of undrained is governed during tunnel excavation,
Therefore the contracted volume at tunnel's span, equals tovolume of clay soil on ground surface. The amount of name it,
V L follows as eq.(2)
4Vs
(2)
V =
L

118

HD 2

(7)
g = GP + U 3 D + ω
In which the difference between maximum internal diameterof excavated tunnel and external diameter of tunnel lining isfor circular tunnel. Equivalent 3D Deflections are, in form ofelastic-plastic at the front of the tunnel and determinant of
excavation quality and operation of the tunnel.
2. TUNNEL GEOMETRY
Fig 3 shows complete sections of twin tunnel of Isfahan
subway. It is a two-arc tunnel that .

Fig 2: Curve of ground surface settlement due to tunnel
excavation [12]Where D is diameter of tunnel and V L is reduction of groundvolume relative to excavation. V L was calculated as apercentage of settlement of volume in ground surface totunnel volume per unit of length. Vs is the total volume ofground surface settlement and is determined by integrating
the relationship eq.(1) as in eq.(3):(3)
VS = 2π × IS max
By combining eqs.(2) and (3) and by regarding I = KZ , S max
0.313VLD(4)S =Fig 3. Sample section of the tunnel: (Center of
documentation of urban trains, Isfahan (2003)

2

max

KZo

In this equation, S max is the maximum ground settlement
above tunnel axis, S x ground surface settlement to the
distance of x relative to tunnel symmetry axis and i is thelandmark of the curve.
Although this equation has no theoretical foundation, it has
been chosen due to its similarity to experimented values ofsettlement. As it was seen, i is an important parameter inshowing curve width[12].
Sagaseta(1987) [15] presented a solution to determine
stress-strain behavior in an isotropic homogeneousenvironment while the incompact able soil and excavation isnear ground surface. The relation which he proposed forsettlement shape in plain strain state, in equations ( 5) and (6):( )(s xo = − a 2 x / x 2 + h 2)

Horizontal displacement at ground surface

( )(
2

2

s xo = − a h / x + h

2

)

(5)

Settlement at ground surface(6)
In this equations (a) is tunnel's radius, and (h) is the depth oftunnel's axis. Because of incompatibility of soil the Poissoncoefficient is 0.5 presented a linear elastic analysis for atunnel in an isotropic environment. They use elasticity theory
and approximate method of Sagaseta [15]which is evenusable for compactable soils with arbitrary amounts ofPoisson coefficient. Their solution that is the generalized
method Sagaseta, considers oval deflection of tunnel section[16].
Another factor that is defined in different condition of groundsettlement is ground loss parameter that is the percentage ofthe ratio of settlement volume at ground surface to tunnel
volume per unit of length. Based on ground loss, gap
parameter is defined as follows:

Based on the first stage research, urban train of Isfahan was
designed in the form two dimensional and horse-shoe section,
7m in the middle part and its axis to axis distance is 14m.
Fig 5. is a part of target vertical section in the direction oftunnel excavation. This section was built in the width of
Chahar Bagh street. Due to symmetry of boundary andenvironmental condition of these two tunnels, only half ofgeometry of target section, regarding environmental
condition and loading shallow adjacent to tunnel is shown[10].
This street has two driving lines and 7m of its middle section
is occupied by landscaping . The street width is assumed to be
90m in which 22.5m from each direction, is allocated to thebuildings. Loading system of building is accounted 20 kP in
two ways and loading of middle part (traffic loading) is
supposed to be 10 KP [11] .

Fig 4: Vertical section of the symmetrical geometry of thetunnels and the region under study

119

So in the first stage, regarding the present situation, thegeometry of modeled section is made in the form of two
dimensional by PLAXIS software and based on Fig 5.

(a)

Fig 5: Modeled geometry of tunnels in software space for5 meter depth [10]3. DETERMINING THE SPECIFIC FEATURES OF
GEO TECHNICAL MODEL
Based on present reality and environmental condition, in thismodel, geology layers and behavioral model of soil are
assumed horizontal and coarse-grained respectively. Theselayers are defined in three different parts and in the form of
fill materials, alluvial and silt from up to down. Geotechnicalparameters given in Table 1.

Fig6: Comparison of shallow settlement profile for six
states of tunnel depth (a) - The varying trend of maximum
settlement relative to excavation depth (b)[10]

Table 1: Behavioral model and geotechnical specificationof soil [11]FillMaterial

Alluvial

Shale

Unit

SymbolParameter

0-2.5

2.5-21.5

21.5-30

m

H

18

21.5

26

kN/m3

γd

5*10-7

2.5*10-7

5*10-6m/s

Kh

1*10-6

2.5*10-4

1*10-6m/s

Kv

1.5*104

8*104

1.7*106

kN/m3

E

0.35

0.3

0.3

-

ν

Depth
Unit
WeightPermeabilityPermeabilityYoungModulus
PoissonRatio
CohesionInternal
Friction
Angle

3

15

50

0.8

kN/m

C

23

35

26

Degreeϕ(b)

4. ANALYSIS OF OUTPUT DATA
Based on carried out modeling in four different andsubsequent phases, the results for shallow settlement, for six
different states at tunnel depth are achieved. According tothese results, it is clear that maximum settlement in groundsurface is occurred around the tunnel axis in all states, and byincreasing the surcharge of the tunnel, the interference ofsettlement depth of two tunnels is also increasing to the pointthat in surcharge state of 10m , a ditch of settlement is seen
completely. The varying trend of maximum settlement
relative to excavation depth of tunnel is shown in Fig 6. Ii isobvious that by increasing the depth of tunnel excavation, theamount of maximum settlement is significant [10] .

The results show that:a- In different states of surcharges(5 up to 10 meters),
settlement in neighborhood of buildings in bothsides of street is within the allowable of 1.0
centimeter.
b- Maximum settlement at ground surface happened atneighboring points of axis of tunnel.
c- By increasing the height of rock load on tunnels, theinteraction of ditches of resulted settlement us
increased.
d- the amount and area of settlement of ground surface,has direct relationship with depth of tunnel
excavation.
e- It must be noted that the modeling can be done byexamining three-dimensional effects of front ofactivities
on shallow settlement amounts inneighborhood of tunnels and assess the results [10]5. EFFECTS OF THE DISTANCE BETWEENPARALLEL TUNNELS ON SETTLEMENT LEVEL
Instrumentation studies show that if center to centre distanceof two underground space is more than three times or much as
diameter, the excavation of these two spaces will not affecteach other, and can be considered separately. In this section
we will address this issue by numerical method andexamining a tunnel.

120

First, the settlement from a tunnel excavation in an urban
environment is modeled, and then the effect of excavating
another tunnel in its neighborhood on the settlement is
considered. A tunnel with 4.5m diameter and 10m depth isexcavated. Before tunnel excavation the present stress inenvironment is in the horizontal form. It should be kept inmind that the best and optional distance for neighboring
borders is the amount that, stresses at borders, has notchanged just before and after the excavation and disturbancein stress contour doesn't happen[fig7][Fig8][Fig9] .
Fig10: Total strain - single tunnel

Fig7: Initial stresses before excavating the tunnel

Fig11: Total stress - twin tunnel with 2m distance

Fig8: Stress state after excavating 10m of the tunnel

Fig12: Total strain - twin tunnel with 2m distance

Fig9: Single tunnel total displacement

After that the effect of neighboring tunnel excavation isconsidered with the same characteristic and in different
distances. To do this, 5 states have been regarded. The
distance we mean in this study is the ledge to ledge distanceof two neighboring tunnels. The purpose of examining
several cases (states) is obtaining optimal distance between
two neighboring tunnels. At first neighboring (adjacent)tunnel is put in two meters distance from the first tunnel [Fig10][Fig11][Fig12] .
As it can be seen, stress contours have overlap and do not act
independently so that stress contour of twin tunnel is notsimilar to that of single tunnel, and this is also true for soil
displacement around the tunnel. In the following situation
tunnels are placed in 4 meter distance. In this case, although
the interaction of stress and displacements contours in twotunnels are decreasing, but they still influence on each other.
121

In another situation, two tunnels are put in 8m distance fromeach other. It may be said that behavior of two tunnels ispartially independent and have the least effect on each other[Fig13][Fig14] .

Fig16: Total strain- twin tunnel with 10 m distanceFig 13: Total strain twin tunnel with 8m distanceFig 17: Total stress- twin tunnel with 12m distanceFig 14: Total stress with tunnel with 8m distanceIn another situation two tunnels are put in 10m distance and as
expected, by increasing this distance and keeping the tunnels
away from each other, their behavior becomes more separate[Fig15][Fig16] .

Fig 18: Total strain -twin tunnel with 12m distanceFig 15: Total stress- twin tunnel with 10m distanceIn the next situation two tunnels are put in 12m distance. As itwas seen from the distance of 10m or more, stress contours
and displacement around each tunnel are just alike single
tunnel and next tunnel excavation has no effect on
displacement and applied stresses on neighboring tunnel sothat to simplify and reduce the calculation, we can examine
them independently (separately) [Fig17][Fig18] .
6. COMPARING OF SINGLE TUNNEL BEHAVIOR
WITH PARALLEL TUNNEL (WITH EQUALEFFICIENCY)
In this part a tunnel with 9 m diameter (two times as muchdiameter as former tunnel) is excavated and studied at thesame environment and condition. The purpose of this study is122to examine settlement from tunnel excavation with larger
diameter and its comparison with settlement from twotunnels, each one with a diameter equals to the radius oflarger tunnel (with different distance from each other)[Fig19][Fig20] .

Fig 19: Total stress-single tunnel

d=9m

Fig 20: Total strains-single tunnel

d=9m

settlement were drawn and the result of created effect on thesettlement of ground surface and tunnel crown was studied
due to increasing or reduction of each of the mentioned
parameters. Finally by modeling a tunnel with 9m diameteron one hand and two neighboring tunnels, each one with 4.5m
on the other hand, and putting them in different distance fromeach other and observing created changes in stress contour
and displacement as to using or not using two tunnel withsmaller diameter instead of a tunnel with larger diameter inurban regions were explained. Regarding above –mentioned
materials, following results are concluded:
1. The result from in situ measurement indicates thatmaximum settlement from excavation of subway tunnel inIsfahan in monitored regions is less than amount of allowablesettlement (0.5 under the building as 1 inch at street level).
2. based on done research on stress contours and length
displacement of 9 m diameter, and its comparison with stresscontours and displacement of a twin tunnel-each with 4.5 m
diameter at different distance from each other and withsimilar and same condition of a larger tunnel, we may say thatin situations where tunnel construction is in urban regions orother sensitive places(areas), naturally any increase will havesome important consequences. Using twin tunnels withsmaller diameter seems reasonable and justified if theallowable distance for each of the tunnels-instead of
excavating a tunnel with larger diameter- is observed.
3. In cases where the distance of two neighboring tunnels areequal or more than allowable level, we can model one of twotunnels and use results for another tunnel and generalize its
result(s) to another tunnel.
8. REFERENCES

As it was seen in comparing the displacement fromexcavation of two neighboring tunnels (with smaller
diameter) and larger tunnel, it would be better to excavate twotunnels separately so that their distance should be inallowable level and has not much effect on displacement andsettlement of ground surface.
7. CONCLUSION
In this study first the situation (condition) of urban train
tunnels is studied in Isfahan and then a comprehensive review
of literature is discussed about created deflections on anunderground excavation and their settlement. Then general
overview of stability analysis approaches are considered andregarding the use of PLAXIS software some brief description
is provided. Two cross sections of subway tunnel of Isfahan
were studied in which the settlement of ground surface andtunnel crown were measured by instrumentation equipments
and was modeled by PLAXIS 3D. The outputs of software
were adapted with measured data in above supposed stations.
Furthermore, in order to examine the effect of importantparameters of soil (such as cohesion, Poisson coefficient,elasticity modulus and unit weight) different diagrams of[1]- Terzaghi,k. Rock defects and loads on yunnel supports .
in: Rock tunneling with steel supports.editors r.v.proctor andt.white . Published by commercial shearing and stamping co .
, Youngstown,1946,pages 15-99[2]- Shariatmadari,N, Farzaneh, (2001), investing settlementof ground surface due to tunnel excavation in Tehran. p.p
265-277 the proceedings of 5th conference of Iran's tunnel
Tehran's Technical- Department of Tehran University[3]- Vafa'eeyan, M (2001): the characteristics of groundsettlement owing to shallow tunnel excavation p.p 313-321[4]- Proceeding of 5th conference on Iran:Tunnel Tehran
Technical engineer Department of Tehran University The
collection of reports of document center of subwayorganization of Isfahan and neighbor hood (2003), Report ofanalysis and design of maintenance system and installing
program of exact instrument and of first section.
[5]- The collection of documentation center reports ofsubway organization of Isfahan and neighborhood (2008):
monthly report of instrumentation of subway tunnel ofIsfahan, southern piece (part).
[6]- The company of consulting engineers of Zamin Fan
Avaran (2005). Final report of geo technical studies southernpart of Isfahan metro subway.

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[7]- Collections of reports of documentation center of subwayorganization of Isfahan and neighborhood (2001) potentialityof fast and dense transportation of Isfahan[8]-Raftari, P. (2011) Analysis of stress and deflections of thetunnel of Isfahan subway by using the results of
instrumentation and its adaption with results gained by
PLAXIS 3D tunnel software, M.Sc thesis Azad University of
Zanjan Engineering Department.
[9]- Khosrotash Mehran, Taskindoust, kashfi mojtaba (2007).
Detailed report of Behavior Assessment tunnel, first section
2nd line karaj and neighborhood subway consulting engineersof Rad Tunnel No 4.
[10]- Majid Nourian bidgoli, (2009): Studying the settlementof ground surface due to excavation of twin tunnels in urban
areas 8th conference of Tunnel. Tarbiat Modares,
Tehran(2004)[11]-Consultant engineers of Saazbon porozhe,.reports offirst stage study of geology and geotechnical of Isfahan
subway within the Shariati-Azadi Station subwayorganization of Isfahan and neighborhood, (2004).
[12]- Peck, R. B. (1969). "Deep excavation and tunneling insoft ground." Proceeding of 7th int. Conf. on soil mechanic,
Mexico: state of the art, pp. 225-290.
[13]- new& O'Reilly. (1982) "Representing Settlement forSoft Ground Tunnleing" Journal of Geotechnical andGeoenviromental Eng., Vol. 124, No.9, pp. 846-856.
[14]- Mair, R. J., Tailor, R. N., and Bracegirdle, A. (1987).
"Surface settlement profiles above tunnels in clay." Journal of
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[15]- Sagaseta, C. (1987). "Analysis of underground soil
deformation due to ground loss". Journal of Geotechnique,Vol. 37, No. 3, pp. 301-330.
[16]- Verruijt, A. and Booker, J. R. (1996). "Surface
settlement due to deformation of a tunnel in an elastic halfplane." Journal of Geotechnique, Vol. 46, No. 4, pp. 753-756.
[17]- Rowe, R. K. and Kack, G. J. (1983). "A theoricalexamination of the settlement induced bytunneling."Canadian Geotechnical Journal., Vol. 29, pp.
299-314.
[18]- Lee, K. N., Rowe, R. K. and Lo, K. Y. (1992).
"Subsidence owing to tunneling, Estimation the gap
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929-940.
[19]- Najjar, Y. and, M. (1993). "Surface subsidence
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[20]- Selby, A. R. (1999). "Tunnelling in soil-groundmovements and damage to building in Workington, UK.
"Geotecnicl and Geological Engineering, Vol. 17, pp.
351-371.
[21]- Wang, Z. W., Sampaco, K. L., Fischer, G. R., Kuchker,
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for predicting surface settlements due to soft groundtunneling. Boston: Proceeding of North American
Tunnelling, pp. 645-652.

124

125

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Excess Pore Pressure Characteristics of Sand Mat using Dredged Soil
M.S LEE, K. ODA
Department of Civil Engineering, Graduate School of Engineering, Osaka University, Japan

ABSTRACTThe design of sand mat should be reviewed by behavior of excess pore pressure which is obtained by combining
characteristics of soft ground with the permeability of sand mat. In this paper, in order to investigate the distribution of
hydraulic gradient of sand mat, a banking model test was performed using dredged sand as materials of sand mat, andthese results were compared by the numerical analysis results utilizing Terzaghi's consolidation equation. As theresults, it shows that the pore pressure was influenced by the settlement increasing in the center area of sand mat as theheight of embankment increases, and uprising speed of excess pore pressure due to residing water pressure is delayedcomparing with the results of numerical analysis. Finally, the construction of sand mat should be laid to reduce theincreased hydraulic gradient at the center area of embankment.
Keywords: Dredged Soil, Sand Mat, Model Test, Hydraulic gradient, Pressure Head

1. INTRODUCTION
Sand mat is used as a horizontal drainage method during
embankment to facilitate the initial consolidated settlement
due to the weakness of the surface layer of dredged andreclaimed sites, and this construction method is applied ingeneral areas of soft soil ground. In particular, due to theinsufficient amount of sand in dredged and reclaimed areas
and areas of soft soil ground, there is a desperate need to
procure alternative materials, which has increased the needfor research on using quality dredged soil as a sand mat.
Thus, in order to determine whether dredged soil is a suitablematerial to be used as a sand mat, a large sample wascollected for physical properties analysis and permeability
testing. Based on the results, the representative dredged soil(of the samples with permeability coefficients that fit thecriteria for sand mat, the sample with the worst particle size as
it contains the most amount of fine powder) was selected anda model test was conducted to determine its applicability as asand mat in advance. The distribution of pore water pressure
and the drainage of the sand mat, which is necessary in the
improvement of soft soil ground, were identified and theamount of consolidated settlement according to theembankment load was measured in order to perform afinite-element analysis and predict its behavior. After the
consolidation was completed by the final load, a discharge
test was performed and the effect of the permeability wasobserved. The aim of this study is to apply the theoreticalreview to the design by proposing a testing technique that canbe utilized on sand mat, and improving the technology ofutilizing alternative materials for sand mat based on the
results.

2. THEORETICAL WATER PRESSURE
DISTRIBUTION IN SAND MAT
The pore water pressure of sand mat laid on the soft clay
ground generally increases in proportion to the amount ofsettlement in the clay layer according to Terzaghi's theory of
one-dimensional consolidation (Terzaghi, 1943, Terzaghi, etal., 1967). The increased excess pore water pressure isquickly drained out from the embankment body due to thepermeability of the sand mat. However, as shown in Fig. 1, if
the permeability of the sand mat is low or there is large
settlement of the soft clay ground, drainage is delayed in thecentral part of the embankment, which increases the pressure
head. In Fig. 1, it is assumed that the amount of consolidated
settlement in the saturated clay layer generated by theembankment load undergoes a one-dimensional change andthe pore water from consolidation is completely drained out
from the embankment body in a horizontal direction throughthe saturated sand mat. Thus, for saturated clay ground, theconsolidated settlement of the clay layer can be represented asa function of the average degree of consolidation (U) and time(t) according to Terzaghi's consolidation equation. The
pressure head (H) occurring at the horizontal distance (x) of
the sand mat can be presented as Eq (1) according toTerzaghi's one-dimensional consolidation equation.
L

H0t
Pressure Head(H) CurveHt
Tsand matment

y Settle
Arbitrar

Actual Settlement

tlement
Final Set

Fig.1 Distribution of pressure head of sand mat

126

H=

S 0 f − S Df 2 
γ w dU (t ) ⋅⋅  S 0 f ⋅ x −⋅ x dx (1)dt 
2k ⋅ T
2L


where dU(t)/dt is the settlement rate of the clay layer, S 0f is
the final amount of settlement in the center of theembankment, S Df is the final amount of settlement in the sidesof the embankment, T is the thickness of the sand mat, k is thepermeability coefficient of the sand mat, 2L is the full widthof the sad mat, and γ w is the unit weight of water. Because
Eq (1) assumes a one-dimensional, horizontal flow of thepore water from the sand mat, the pressure head is zero (H=0)
at the sand mat boundary (x= L). Thus, when the maximum
pressure head at the center of the embankment is set as H 0t
and the average pressure head in the transverse distribution isindicated as H t , the pressure heads can be presented as follow
in Eq (2) and Eq (3).

H 0t =

γ w  S 0 f S Df  2 dU (t )L ⋅+2k ⋅ H  3
6 
dt

(2)

L

Ht =

∫ h ⋅ dx0Lγ w ⋅ L2  5S 0 f S Df  dU (t )=+
2k ⋅ T  24
8  dt(3)

In Eq (2) and Eq (3), the pressure heads of the sand mat varydepending on the settlement rate, final amount of settlement,
sand mat thickness, permeability coefficient, and the width ofthe embankment. Here, the settlement rate can be obtained
from the approximation formula of the relationship between
the degree of consolidation (U) and time factor (T v ) in
Terzaghi's one-dimensional consolidation equation and themeasured settlement as follow in Eq (4) and Eq (5).

Sf =

Cc
P + ∆p× H × log 0
1 + e0P0

U (%) =

4

π

Tv =

4

π

×

(4)

Cv ⋅ tS= t × 100
2
Sf
H

(5)

In Eq (4) and Eq (5), S f is the final amount of settlement,
whereas S t is the amount of settlement at an arbitrary time
and the other coefficients have the same symbols as those in
Terzaghi's one-dimensional consolidation equation.
3. MODEL TEST ON BEHAVIOR OF PORE WATERPRESSURE IN SAND MAT
3.1 Experimental Apparatus and Method
For the model test, head devices were designed to be placedon both sides of the soil tank (dimensions: 300cm × 70cm ×70cm) to maintain the groundwater inside the sand mat at aconstant level, and 4 pore water pressure cells, 3 settlement
plates and 2 earth pressure cells were installed to measure theamount of settlement of the clay layer and sand mat and thepore water pressure during embankment loading. The soft
clay in the bottom of the sand mat was sufficiently stirredwith distilled water using an electric agitator in a simple soil
tank, and the slurry was inserted up to a height of 30cm. Then,saturated water was supplied for approximately a week by thehead devices until a steady state was reached. Later, a sheet
was laid, upon which dredged soil was spread to a height of
20cm in place of the sand mat. The drainage valves on the left
and right sides (Fig. 2) were closed and saturated water was
supplied for a day. Then, spreading and loading were
performed up to the third level using weathered soil.
3.2 Physical and Mechanical Properties of Sand Mat and ClayThe physical properties and particle distribution of dredged
soil used in this experiment are shown in Table 1 and Fig. 3.
Fig. 3 also contains information on the general particle
distribution of sand mat applied in general soft soil grounddesigns. The dredged soil used in this experiment is slightly
poor compared to quality sand as it is close to the lower limitof the fine materials used for sand mat, but it is deemed tohave a good alternative to other materials for sand mat. In
order to examine the consolidation characteristics of thelower clay used in this experiment, overconsolidated clay was
turned in a slurry form using an agitator. Then, theconsolidation sample was produced using a consolidation
device. The standard consolidation characteristics of the test
sample and the input moisture content (w n ), etc., of the slurryform inserted into the model soil tank are shown in Table 2.

Dial Gauge

Step 3 ofEmbankment
Water Pressure Cell
Earth Pressure Cell
Settlement Plate

Step 2 ofEmbankment
Step 1 ofEmbankment

Drain Valve
Sand Mat
Clay

Fig.2 Model test on hydraulic gradient of sand mat (unit: cm)

127
Table.1 Physical properties of sand mat used in experiment
Water Content (%)
17.6
Plasticity Index (I p )NPSpecific Gravity (G s )2.60
Passing 75 μm sieve (%)
9.5~13.3
Coefficient of Permeability (cm/sec)

1.34 x 10-3

USCS

SM

4 NUMERICAL ANALYSIS OF PORE WATERPRESSURE OF SAND MAT

100

Criteria of Sand Mat
Dredged Soil

Percentage(%)

80

60

40

20

0
10

1

0.1

0.01

Grain Size(mm)Fig.3 Distribution of the pressure head of a sand mat
Table.2 Physical and mechanical properties of remolded clay
Water Content (w n , %)
54.5
16.6
Unit Weight (γ t , kN/m3)
Specific Gravity (G s )
2.60
e0
1.26
Cc
0.380
Cs
0.159
Consolidation

USCS

Table.3 Amount of settlement and ground stress measured
during embankment
S Df (cm) S 0f (cm) Δp (kPa)
Sand Mat
0.89
1.33
3.4
Step 1
1.35
1.87
3.6
Embankment Step 2
1.76
2.32
5.1
Step 3
2.04
2.62
5.1

C v (cm2/sec)

1.56 x 10-4

P c (kPa)

10
CL

3.3 Settlement Characteristics According to
Embankment
For the embankment, the degree of consolidation was
confirmed to be over 90% using the hyperbolic method (Tan,et al., 1991) and loading was performed in 3 steps up to aheight of 100cm over the course of 11 days. The degree ofconsolidation according to the settlement of the clay ground
was assessed using the ratio of the amount of settlement
actually measured through the model test and the final
amount of settlement calculated using the specifications of
the consolidation test. The final amount of settlement wascalculated based on the initial void ratio according to the
input moisture content and the ground stress (Δp) measured
using the earth pressure cell at the bottom of the clay layer.
The measured values (Δp) of the earth pressure cell at thecenter, the amount of settlement (S 0f and S Df ) measured using
the dial gauge at the center and the left and right fore-ends inthe model test device (Fig. 2) during the embankment are
shown in Table 3.

4.1 Pore Water Pressure According to Permeability
Coefficient and Thickness of Sand Mat
The settlement rate (dU(t)/dt) for the initial consolidation
degree (U) of 3%, obtained when the clay layer thickness of
30cm and consolidation coefficient in Table 2, are substituted
into Eq (5) and the amount of settlement at the center of theembankment (S 0f ) and the amount of settlement in the outerpart of the embankment (S Df ) measured at step 3 ofembankment, and the spread thickness (T) of 20cm, were
substituted into Eq (2). When the permeability, k, changed up
to 10-4～10-2 (cm/sec), the pressure head, H t , was calculated
according to the horizontal distance of the sand mat, and theresults are shown in Fig. 4. As shown in the figure, the lower
the permeability coefficient of the sand mat, the higher thehydraulic gradient at the center of the embankment, which inturn increases the pressure head. In particular, there is asignificant increase in the pressure head when thepermeability coefficient of the sand mat is below 10-4
(cm/sec) as it causes a delay in the draining of pore water.
Also, the settlement rate calculated when the degree ofconsolidation was 3% using Eq (2), the amount of settlement
at the center of the embankment (S 0f ) and the amount ofsettlement in the outer part of the embankment (S Df )measured at step 3 as shown in Table 3 were substituted into
the equation, with the permeability coefficient, k, maintained
at a constant of 10-3 (cm/sec). Then the thickness of the sandmat (T) was varied in the range of 5 to 30cm to analyze theresulting value of the pressure head, the results of which are
shown in Fig. 5. The figure shows that with an increase in the
thickness of the sand mat, the pressure head decreased, andthe change in the pressure head was small in the outer part ofembankment as it is less affected by the hydraulic gradient.
4.2 Pore Water Pressure of Sand Mat Based on
Coefficient of Consolidation in Clay Layer
In order to analyze the pressure head of the sand mataccording to the settlement rate of clay in the bottom, thecoefficient of consolidation (C v ) was selected as a factor
affecting the settlement rate in the clay layer. when the
consolidation coefficient (C v ) of the clay layer is changed up
to the range of 10-5～10-3(cm2/sec) with the initial degree ofconsolidation (U) at 3%, the consolidation time (t) for thearbitrary consolidation coefficient can be calculated using Eq(5) and thus, the settlement rate (dU(t)/dt) with respect to thedegree of consolidation can be obtained.

128
0.8

3.5

Pressure Head Ht (cm)

3.0
2.5
2.0
1.5
1.0
0.5

Cv=10-3cm2/sec
Cv=10-4cm2/sec
Cv=10-5cm2/sec
T = 20cm

0.7

Pressure Head Ht (cm)k=10-4cm/sec
k=10-3cm/sec
k=10-2cm/sec
U = 3%
T = 20cm

0.6
0.5
0.4
0.3
0.2
0.1
0.0

0.0

T

Sand Mat
-91

-60

-30

0

30

60

-30

-60

0

30

60

91

Horizontal Distance (cm)Horizontal Distance (cm)Fig.4 Pressure head according to changes in permeability
coefficientFig.6 Pressure head according to changes in consolidation
coefficient
1.5

T = 5cm
T = 10cm
T = 20cm
T = 30cm
U = 3%

1.0
0.8

Embankment (m)

1.2

1.0

0.5

0.4
0.2
0.0

T

Sand Mat
-91

-60

-30

0

30

60

Settlement (cm)

0.6
0.0
0.2 0
0.4
0.6
0.8
1.0
1.2
1.4
1.6

Elapsed Time (min)
2000

4000

6000

8000 10000 12000 14000 16000 18000

Left
Middle
Right
FEM

91

Horizontal Distance (cm)Fig.5 Pressure head according to changes in thickness of sandmatThus, the amount of settlement at the center of theembankment (S 0f ) and the amount of settlement in the outerpart of the embankment (S Df ) measured at step 3 as shown in
Table 3 and the permeability coefficient, k, a constant of 10-3
(cm/sec), were substituted into Eq (2) to obtain the pore waterpressure in the sand mat. The results are shown in Fig 6,
which shows a positive correlation between the consolidation
coefficient of the clay layer and the pressure head at the centerof the embankment. This is due to the increase in the amountof consolidation drainage resulting from the increased
settlement rate of the clay layer caused by increased stress at
the center of the embankment.
5 ANALYSIS AND DISCUSSION OF RESULTS
5.1 Pore Water Pressure of Sand Mat According toHeight of Embankment
The top of Fig 7 shows the results of the FEM analysis
conducted based on the modified cam-clay model (Snadhu, etal., 1969, Mayhe, 1980, Gens, 1988) of the settlement,
measured in the settlement gauges installed in the left, central
and right parts of the sand mat layer, according to the elapsed
time during embankment. The bottom of the figure shows themeasurements of the pore water pressure cells installed inside
the sand mat layer and at the center of the clay layer.

Pore pressure (kPa)

Pressure Head Ht (cm)T

Sand Mat
-91

91

5

Sand-Mat
Left
MiddleRight
Clay
middle

4
3
2
1

0
0

2000

4000

6000

8000

10000 12000 14000 16000 18000

Elapsed time (min)Fig.7 Pore water pressure of sand mat and settlement and porewater pressure of clay groundAs the height of the embankment increased, there was ahigher amount of settlement in the central part of the clay
layer compared to the left side. Because of the soft surface
and the low shear strength, there was a immediate settlement
during the laying of the sand mat and the primaryembankment, and the actual measurement of the settlement
was higher than the settlement calculated based on FEM
analysis. Afterward, the measured settlement and the result of
the FEM analysis tended to be similar, but the actual
settlement was slightly smaller due to the buoyancy effectcaused by inhibited drainage during the experiment. Also, acomparison of the pore water pressure in the clay layer andsand mat layer during embankment showed a significant
increase of pore water pressure in the clay layer, whereas suchincrease was prominent at the center and low in the left andright sides of the sand mat layer. This is due to the increased
hydraulic gradient caused by an increase in the residual porewater pressure resulting from the concentrated stress in thecentral part of the clay layer.

129
1.8

1.8

1.6

Left
Middle
Right

1.4
1.2

Pore pressure (kPa)

Pore pressure (kPa)

1.6

1.0
0.8
0.6
0.4

1.2
1.0
0.8
0.6
0.4
0.2

0.2

0.0

0.0
200

400

600

800

1000

0

1200

2000

4000

6000

8000

Time (min)Time (min)Fig.8 Changes in pore water pressure during laying of sand mat

Fig.10 Changes in pore water pressure during step 2 of embankment

0.9

3.0

0.8

Left
Middle
Right

0.7
0.6

Left
MiddleRight

2.5

Pore pressure (kPa)

Pore pressure (kPa)

Left
Middle
Right

1.4

0.5
0.4
0.3
0.2

2.0

1.5

1.0

0.5

0.1
0.0

0.0
0

200

400

600

800

1000

1200

1400

1600

Time (min)

0

1000

2000

3000

4000

5000

Time (min)Fig.9 Changes in pore water pressure during step 1 of embankment

Fig.11 Changes in pore water pressure during step 3 of embankment

Fig. 8 to 11 show the results of the pore water pressure
measured using the pore water pressure cells installed in thesand mat and the clay layer according to the embankment
during the model test. Fig. 8 shows a rapid increase of porewater pressure caused by soil disturbance and thegroundwater level remaining at the top of the saturated soft
clay layer. The pore water pressure increased up to 1.6kPa at
the center of the sand mat, and gradually decreased over time
until it reached 0.6~0.4kPa after 400 min. Fig. 9 shows theresult of the pore water pressure measured during the primaryembankment, which slightly increased at the center to a
maximum of 0.8kPa and greatly increased in the left and rightsides to a maximum of 0.6kPa in comparison with the loadingof the sand mat. Fig. 10 shows the pore water pressure
measured during the step 2 of embankment. The pore waterpressure at the center increased up to 1.6kPa at initial loading
before decreasing to approximately 0.5kPa over time. The
bottom clay foundation subsided uniformly until the step 2 ofembankment, which showed that the sand mat has maintained
the horizontal function. The changes in the pore waterpressure of the sand mat at the time of final embankment
shown in Fig. 11 revealed that there was a significant increaseof pore water pressure in the central part of up to 2.6kPa andin the left and right sides of up to 2kPa during initial loading
compared to the step 2 of loading.

Over time, the pore water pressure in the central part of thesand mat was observed to be higher than the left and rightsides, with a constant slope. There was an increase of
settlement at the center of the embankment and the sand mat
formed a concave shape, which caused the pore water to form
a pool as it could not be drained out.
5.2 Pressure Head of Sand Mat According to Degree ofConsolidation in Clay Layer
In order to compare and identify the tendency of the waterhead of the sand mat according to the degree of consolidationin the clay layer during the embankment based on theory andexperiment, the pressure head measured using the pore waterpressure cells installed in the sand mat and the pressure headpredicted using Eq (2) are indicated in Fig 12 and 13. The
thickness and consolidation coefficient of the clay layer in theexperiment were substituted into Eq (5) to obtain thesettlement rate (dU(t)/dt) with respect to the predetermined
degree of consolidation, which was then substituted into Eq(2) to calculate the pressure head. Then, the degree ofconsolidation of the clay ground was obtained using the ratioof the amount of settlement (S t ) measured through the model
test and the final amount of settlement (S f ) calculated using
Eq (4). The pore water pressures of the laboratory clay groundcorresponding to the predetermined degree of consolidation
are indicated in Fig 12 and 13.

130
2.0

1.6
Calculated Measured

Pressure Head Ht (cm)

1.2
1.0
0.8
0.6
0.4
0.2

Calculated Measured

1.8

U=4.5%
U=13.8%
U=25.8%
U=40.8%
U=63.1%

1.6

Pressure Head Ht (cm)

U=4.5%
U=13.8%
U=25.8%
U=40.8%
U=63.1%

1.4

1.4
1.2
1.0
0.8
0.6
0.4
0.2
0.0

0.0

Sand Mat

Sand Mat

Fig.12 Comparison of results and numerical analysis of porewater pressure according to degree of consolidation
during step 2 of embankment

Fig.13 Comparison of results and numerical analysis of porewater pressure according to degree of consolidation
during step 2 of embankment

An analysis of the step 2 and step 3 of embankment, whichcaused a significant increase in the pore water pressure,showed that the pore water pressure of the sand mat measured
during the step 2 of embankment in the model test was
generally higher due to the increase in the degree ofconsolidation compared to the result of the numericalanalysis. The measured pore water pressure was at its
maximum at the center when the degree of consolidation was
26% and at the left and right sides when the degree ofconsolidation was 14%. On the other hand, the result of thenumerical analysis revealed that the maximum pore waterpressure was reached when the degree of consolidation was
5%. Fig. 13 shows that the pore water pressure was measured
to be at its maximum at the center during the step 3 ofembankment when the degree of consolidation was 14% andat the left and right sides when the degree of consolidation
was 5% and 14%. On the other hand, the result of thenumerical analysis revealed that the maximum pore waterpressure was reached when the degree of consolidation was
5%, just as with the step 2 of embankment. In the case of step
3 of embankment where the height of the embankment isgreater, the pore water pressures measured in the left and rightsides were higher compared to the values measured during thestep 2 of embankment, and with an increase in the degree ofconsolidation, the measured values for the left and right sidesof the sand mat were greater than the results of the numericalanalysis. As the embankment grew taller, there was anincrease of settlement in the central part of the sand mat,causing a deformation, which inhibited drainage. Thus,(1) Terzaghi's one-dimensional consolidation equation,
which takes into account the rate of consolidated settlement in
the clay layer and the settlement level, is highly influenced by
the permeability coefficient and thickness of the sand mat andthe consolidation coefficient of the clay layer.
(2) Because the settlement level of the embankment center
increases pressure head, it reduces the horizontal drainage ofsand mat.
(3) The reason for the delay in the drainage of the pressure
head measured during the test compared to the result of thenumerical analysis is the increased hydraulic gradientoccurring because the increased settlement in the central partis transferred to the left and right sides.
(4) The amount of settlement in the central, left and rightsides should be predicted prior to the laying of the sand mat inorder to reduce the hydraulic gradient and the horizontal
water head resistance in the sand mat.

6 CONCLUSION
In this study, the pressure head distribution of a sand mat laid
on a clay layer was analyzed based on Terzaghi's one dimensional consolidation, which takes into consideration thesettlement characteristics of the bottom clay layer. Also, theeffect of the settlement characteristics of the soft clay layer on
the pore water pressure distribution of the upper sand mat
layer was reviewed based on the model loading test. The
findings of this study can be summarized as follows:

7 REFERENCES[1] Terzaghi, K, “Theoretical Soil Mechanics,” New York:John Wiley and Sons, 1943.
[2] Terzaghi, K. & R. B. Peck, “Soil mechanics inengineering practice,” 2nd ed. New York: John Wileyand Sons, 1967
[3] Tan T. S., Inoue. T., Lee. S. L, “Hyperbolic method forconsolidation analysis,” Journal of GeotechnicalEngineering, vol. 117, No.11, 1991, pp.1723-1737.
[4] Snadhu, R, S. & E. L. Wilson, “Finite Element Analysis
of Seepage in Elastic Media,” J. Engrg. Mech. Division
ASCE, Vol.95, No. EM3. 1969, pp.641∼651.
[5] Mayhe, PW, “Cam-Clay predictions of undrained
strength,” Journal of the grotechnical andgeoenvironmental engineering, ASCE, Vol.106, No.
GT11, 1980, pp.1219∼1242.
[6] A. Gens, D.M. Potts, “Critical state models incomputational geomechanics,” Engineering
Computations, Vol. 5 Iss: 3, 1988, pp.178 – 197.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect Of Clays Fraction To California Bearing Ratio Laboratory Test Value
With And Without Soaked
Soewignjo Agus NUGROHO, Syawal SATIBI and Ferry FATNANTA
Civil Engineering Department, Riau University, Indonesia

ABSTRACTIn design of highway, soil bearing capacity is very affecting the thickness of pavement. One of the methods to determine
the soil bearing capacity in Indonesia is CBR (California Bearing Ratio) test. The field soil conditions are soaked(saturated) and un-soaked, considering these conditions, conducted CBR testing in the laboratory on soaked andun-soaked conditions.
This research aims to make comparisons between CBR soaked test results for CBR un-soaked in some variation of clay
content and make simple comparisons between CBR soaked for CBR un-soaked by considering the soil properties
whereas can be predicted the CBR soaked value based on the CBR un-soaked test results.
The results showed that there was a linear correlation between the CBR soaked and CBR un-soaked also influenced by
the nature of the index (the properties of the soil). The maximum value of CBR ranges of 30-40 percent clay content.
Keywords : California Bearing Ratio, Soaked and un-soaked, correlation
1. INTRODUCTION
Sub-Grade soil bearing capacity plays very important role
for the design of highway structure. It determines design
thickness of the pavement. High bearing capacity of
sub-grade soils reduces the required thickness of pavement.
The bearing capacity of sub-grade (base soil) is mostly
influenced by the type of soil, water content and its density.
Several methods are available to determine base soil bearing
capacity such as California Bearing Ratio (CBR) test, Plate
Bearing test (to determine modulus of sub-grade reaction andmodulus of resilient), Dynamic Cone Penetrometer (DCP)
test, Machintosh Probe and Hand Cone Penetrometer (HCP)
test, which is also known as Proving Ring Penetrometer [1]It is common in Indonesia that the base soil bearing
capacity for highway pavement design is determined by CBRtest measurement. This can be from the laboratory CBR test
or directly from field CBR test. However, base soil bearing
capacity can also be determined using field tests such as DCP,
HCP, and Machintosh Probe [6]. These tests are much
simpler and faster to perform. Correlation between the resultof DCP test and CBR value, CBR test and HCP test areavailable whereas the prediction between the result of CBRlaboratory soaked test and CBR laboratory soaked value ishardly found. Moreover, this prediction should be determined
locally based on common local experience.

This research is aimed to obtain a local correlation between
the results of CBR laboratory test without soaked and CBRsoaked value. The correlation is based on the comparison
CBR un-soaked test results and CBR soaked value which has
the same fraction of sand and clay in soil

2. LITERATURE RIVIEW
Proctor Compaction Test
The Proctor compaction test is a laboratory method ofexperimentally determining the optimal moisture content at
which a given soil type will become most dense and achieve
its maximum dry density. These laboratory tests generally
consist of compacting soil at known moisture content into acylindrical mould of standard dimensions using a compactive
effort of controlled magnitude. The soil is usually compacted
into the mould to a certain amount of equal layers, each
receiving a number blows from a standard weighted hammer
at a specified height. This process is then repeated for various
moisture contents and the dry densities are determined foreach. The graphical relationship of the dry density to moisture
content is then plotted to establish the compaction curve. The
maximum dry density is finally obtained from the peak pointof the compaction curve and its corresponding moisture
content, also known as the optimal moisture content (OMC).
The testing described is generally consistent with theAmerican Society for Testing and Materials (ASTM)
standards, and are similar to the American Association ofState Highway and Transportation Officials (AASHTO)

131

standards. Currently, the procedures and equipment details
for the standard Proctor compaction test is designated by
ASTM D698 and AASHTO T99. Also, the modified Proctor
compaction test is designated by ASTM D1557 and
AASHTO T180 [2].

performed at nine conditions of sand and clay mixture soil
within the city of Pekanbaru, Indonesia. The result of index
properties can be seen in Table 1. For index properties, soil
can be classification for sand poorly graded (SP) to clay withlow plasticity (CL) [7], [8]2.1 California Bearing Ratio (CBR) test

Table 1 Properties of Materials

The equipment for determining CBR value is a piston
having an area of three square-inches. The piston is moved invertical direction on a soil sample with a speed of 0.05
inch/minute. A Proving ring with dial gauge is attached to thepiston to measure the load at certain penetration. The CBRvalue is the comparison between applied piston loads on a soil
sample and the standard loads, which value is expressed inpercentage [2]).

Clays
0
30
40
45
50
55
60
70
100

Basically, the CBR value describes the strength soil
compared to the standard material. Indirectly, it also
describes the relative density of the soil. Several correlations
between CBR values and the results of other fieldmeasurements exist such as to results of Dynamic Cone
Penetrometer (DCP) test [3], [4], [5]. This has been used inpractice.
A comparative study of HCP and CBR field tests has been
performed by Nugroho et al. [6] on peat, sand, clayey sandand clays soil. The study was aimed to indirectly relate thevalue of CBR un-soaked to CBR soaked value through thecomparison of the results of HCP to CBR field tests. Fromthe point of view of testing mechanism, CBR field test and
CBR laboratory test procedures are the same. CBR field test
uses static penetration whereas HCP is also uses quasi-static
penetration test. Compared to CBR un-soaked test, which isalso a quasi-static penetration test, CBR laboratory is a closer
method. Hence, direct correlation between CBR un-soaked
tests results to CBR soaked value seems to be more relevant.
This correlation can be based on the same soil mixture. This
study aims to obtain direct local correlation between the twolatest two tests.

LL (%)
NP
13,58
16,87
19,16
20,36
20,54
23,81
25,63
37,73

PL (%)
NP
2,43
5,56
6,86
7,84
7,96
9,86
8,32
16,19

Gs
2,671
2,677
2,679
2,681
2,684
2,688
2,704
2,699
2,75

USCS
SP
SM
SM-SC
SM-SC
SC
CL
CL
CL
CL

3.1 Materials
Materials required for field tests are a set field CBR tools
and a CBR mould. The CBR mould was used to obtain
undisturbed sample for determination of physical andmechanical properties of the soil in laboratory and CBRmould uses to determine CBR value after soaked in the
Laboratory for 7 (seven) days. Along mould containing
specimen soaked, swelling of the specimen should be noted toknow swelling potential of the soil from different location.
Figure 1 shows the layout of CBR test in the field (un-soaked)
and in the Laboratory (soaked)

3. MATERIALS AND METHODSIn order to obtain the correlation between CBR soaked
test and CBR un-soaked test results, comparison of two kind
CBR test condition of several mixture soil samples from
Pekanbaru were performed. The CBR tests of two condition
tests were performed for each mix-soil sample from eachvariation. Thus, the density before and after soaked of the soil
for both tests is the same for each soil from each condition.
There were 45 CBR soaked tests and 45 CBR un-soaked tests
132

a. CBR Laboratory

The test results of physical and mechanical properties of thesamples can be divided into four categories based on the typeof soils as seen in Table 2 and Table 3.
Table 2: Physical and Mechanical Properties of CBR
unsoaked Test Samples
Sand (%)

b. Field CBR equipment
Figure 1 Testing Equipment

γd (kN/m3)

CBR (%)

100

0

7.5

16.24

12.710

70

30

10.70

20.50

34.005

60

40

11.75

20.09

14.433

55

45

12.60

19.75

14.738

50

50

12.80

19.65

13.927

45

55

12.85

19.18

13.509

40

60

14.00

19.12

13.411

30

70

14.75

18.68

13.786

0

0

14.85

16.31

4.739

Table 3: Physical and Mechanical Properties of CBR soaked
Test Samples

3.2 Methods
Surcharge weights of 2.5kg are placed on top surface ofsoil. The penetration plunger is brought in contact with thesoil and a load of 4kg (seating load) is applied so that contact
between soil and plunger is established. Then dial readings
are adjusted to zero. Load is applied such that penetration rate
is approximately 1.25mm per minute. Load at penetration of
0.5, 1, 1.5, 2, 2.5, 3, 4, 5, 7.5, 10 and 12.5 mm are noted.
The CBR soaked (CBR tested in laboratory) tests were
performed simply. After mould containing specimen soaked
about 7 days, mould containing specimen is placed in positionon the testing machine, by pressing the piston into the mould
containing specimen from different locations. After that, load
is applied such that penetration rate is 1.25mm per minute andthen the tests were performed. For the determination of thephysical and mechanical properties of the soils, undisturbed
samples were taken from each location and the tests weredone in laboratory.
4. RESULTS AND DISCUSSION
The results of this research are presented in three parts.
First, the results of the all performed tests are described. Afterthat regression analysis between CBR test results and filed
CBR values as well as regression of fraction clay with Gs,
Atterberg limits, unit weight density of the soils are shown. In
the final part, the correlation between CBR soaked tests
results and CBR un-soaked test results are put forward
4.1 Physical and Mechanical Properties of Test Samples

Clay (%) w (%)

Sand (%)

Clay (%) w (%)γd (kN/m3)

CBR (%)

100

0

9.379

16.05

9.573

70

30

10.276

19.13

21.28660

40

12.648

18.18

8.910

55

45

13.726

18.25

4.976

50

50

13.889

18.61

4.979

45

55

14.223

18.16

3.862

40

60

15.510

17.98

3.507

30

70

19.153

17.05

1.625

0

0

24.805

15.71

2.795

For the soils which are considered as in-organic soils (Sand,
Clay, Sand-Clay mixture), in general, they have watercontent, w n between 7.50–14.85%, dry unit weight, γdbetween 16.24–20.50 kN./m3.in field (un-soaked) and havewater content between 9.379-24.805%, dry unit weight
between 15.71-19.13kN/m3 (after soaked in laboratory).
Furthermore, It was recorded that the values of CBR
un-soaked tests on those soils are between 4.739-34.005%
and CBR soaked values between 2.795-21.286% (see Table 2
and Table 3). It is shown that the range of the physical andmechanical properties the soils varies considerably.
As also can be seen in Table 2 and Table 3, for sand soil (sandpoorly graded, SP), its properties (water content, dry unit
weight, CBR value) are not different compared between
soaked and un-soaked condition showing that it has asignificant characteristic compared to the other samples. The

133

soils with clay fraction 70% and sand fraction 30% of sample
CBR value is far above the maximum CBR value of all soils.
The values of its dry unit weight, CBR soaked, and CBR
un-soaked are far below the values of the other soils.
4.2 Regression Analysis of CBR, index properties andunit weight density
Figure 2 to Figure 4 show the results of regression analyses
between clay fraction, index properties and unit weighdensity tests results as well as regression results between
CBR un-soaked and CBR soaked test results. The regression
analyses are made for each clay fraction of soils which are 0%
clay, 0% sand, clay and sand and clay mixture.
It can be seen that regression using linier function suits therelation between water content and clay fraction relatively
accurately whereas for the relation between Atterberg limits
and clay fraction, second order polynomial function shows
relatively accurate approximation. The two regression
analyses will be combined later using Pearson’s correlation
method to find the correlation between CBR test results andAtterberg limit with unit weight density.

Figure 4: Regression Results for Clay
Figure 2 Regression results for Spesific Gravity
Figure 3 shows that the value of liquid limit and plastic limit
of soils increased with increasing clay fraction in the soil.
Gradient increasing of liquid limit higher than plastic limit ofsoil. So, plasticity index of soils increases in proportion to theincrease in soil clay fraction. Figure 3 also shows that
optimum moisture content, determined from compaction test,same as with value of plastic limit at clay fraction between
30% and 70%. It means that for sand clay soils with for 30%
to 70% clay fraction, water content for CBR test can beapproximated by the value of the soil plastic limit.

It can be seen from Figure 4 that for the clay fraction smaller
than 70%, swelling is relatively smaller. Therefore, potential
swelling for soil containing fine grain smaller than 70% is
relatively small.
Figure 4 shows that maximum density is on a combination of
30% clay and 70% sand. In other words, it is the best
composition to obtain the optimal value of CBR is in therange of 30-70% clay fraction. Fraction of clay between 30%
and 70%, soil density is more than 90% of the value of
density of 30% clay and 70% sand.

134

(2)
It was found that value for b 0 b 1 , b 2 , b 3 and b 4 are –25.28,
0.168, -0.064, -2.356 and 3.562 respectively for all soils.
However, for the value of a 3 , there is no unique solution for
all soils

Figure 5: Regression Results for Sand-Clay Mixture
Figure 5 shows that the difference of the results of the CBR
un-soaked and the CBR soaked are increase for 0% to 30%
fraction of clay and then decreases again up to 45% fractionof clay. For clay fraction more than 45% difference between
the CBR un-soaked and CBR soaked tend to a constant.
The correlation formula obtained from the correlation
analyses can be written as follow:

4.3 Correlation CBR Test Results
In the previous section, relations between Clay Fraction andindex properties as well as between clay fraction and soil
density have been obtained. In order to correlate the CBR
un-soaked test results to CBR soaked CBR value, Pearson’s
correlation method is applied to both obtained power andpolynomial functions for each type of soils.

∆CBR [1] =10.78+0.01C+0.01LL-4.41IP+6.37OMC-2.15MDD(3)

On using soil index properties (liquid limit, index plasticity)and the value of unit weight density (OMC, MDD) test as
variables, the following linear equation can be applied to findsimple correlation between CBR soaked and CBR un-soakedon the basis of the same clay fraction value of (Nugroho,
2011)∆CBR [C] =-25.28+0.17C-0.06LL-2.35IP+3.56)MCwith∆CBR [2] =-25.75+0.34C-0.53LL-2.63IP+3.81OMC (4)∆CBR [3] =-25.91+0.09LL-0.98IP+3.80OMC∆CBR [A] =-2.25+2.19LL-2.98IP-0.76OMC

(6)
(7)

It is found that the value of b 0 of -25, suits for equation (4),(5) and (7), clay and sand-clay mixture (in-organic soils)
whereas for the solution of b 0 is for minus 25 to minus 26.
The final correlation formula can be written as follow:∆CBR=-25+C 1 C+C 2 LL+C 3 IP+3.5 OMC

(1)

(5)

(8)

Where C 1 , C 2 and C 3 is 0.34; -0.53 and -2.63 for sand, clay
and sand-clay mixture (in-organic soils). ∆CBR is the valueof CBR un-soaked-CBR soaked.
For soils containing clay between 30 and 70%, the value ofC 2 need to be further tested and the C 3 value is 2.35. The
formula can be rewritten below.

constants= value of CBR un-soaked – CBR soaked (%)= clay fraction (%)= value of liquid limit (%)= value of plasticity index (%)= value of optimum water content (%)∆CBR=-25+0.168 C-0.064 LL – 2.356 IP +3.5 OMC (9)The values of the constants b 0 , b 1 , b 2 , b 3 , b 4 can be solved
using SPSS software which is based on the solution of thefollowing matrix

4.4 Validation of the Correlation Formula135

For the validation of Equation (9), several prediction tests
have been performed. Figure 7a to figure 7b show thecomparison between predicted values of field CBR andmeasured field CBR values for different soil types and soil
densities.

CBR with and without soaked test for in-organic soils (sand,clay and sand-clay mixture). The formula needs to bemodified and further research need to be done for peat soils.
REFERENCES[2] Annual Book of ASTM Standards, 1997, Standard Test
Method for CBR (California Bearing Ratio) of
Laboratory-Compacted Soils, ASTM D1883, WestConshohocken, PA[9] Anonyms, 2007, In Situ Testing and Soil properties
Correlation, International Conference on In Situ
Measurement of Soil Properties and Case Histories, Bali,
Indonesia[7] CRAIG, R., F, 2004, Soil Mechanic, 7th Edition, Taylor &Francis e-library, McGraw-Hill 270 Madison Ave, New
York, USA.
[8] DAS, Braja, M., 2008, Advanced Soils Mechanic, 3rd
Edition, Taylor & Francis Group, McGraw-Hill 270
Madison Ave, New York, USA[1] FARRSHAD A, 2003, Potential Applications of dynamic
and Static Cone Penetrometer in MDOT Pavement
Design and application, Final Report, Department of Civil
Engineering Jackson State University, Corporate
Department of Transportation and US Department ofTransportation Federal Highway Administration,
Mississippi[4] KLIMOCHKO, Dennis, 1991, the Application of the
Dynamic Cone Penetrometer Test in Determining In Situ
Sub-grade CBR. Paper submitted for the 44th Annual
Conference, Saskatoon, Western Association of Canadian
Officials, Geotechnical and Materials Branch,
Saskatchewan Highways and Transportation, Regina,
Saskatchewan, Canada.
[5] SMITH, R.B., and PRATT, D.N.,1983, A Field Study of In
Situ California Bearing Ratio and Dynamic Cone
Penetrometer Testing for Road Design Investigations,
Australian Road Research, pp. 285-294,[3] VAN VUUREN, 1969, Rapid Determination of CBR with
the Portable Dynamic Cone Penetrometer,. Reprint from
"The Rhodesian Engineer".

Figure 6 Predicted and Tested CBR un-soaked (30-70%)

Figure 7 Predicted and Tested CBR soaked (0-70%)
It can be seen from Figure 6 to figure 7; the predicted field
CBR values give significant agreements with the measured
field CBR from the tests for soils that containing clay
between 30% and 70%. On the other hand, very poor
agreements were found for soils contain clay less than 30%
and more than 70%. Hence, the local correlation formula isonly valid for in-clayey sand or sandy clay.
5. CONCLUSIONSThis research has been performed to find local correlation
between CBR un-soaked test results and CBR soaked values.
A linear correlation has been put forward for the local
correlation between the two values. Verification of theformula shows that the correlation can be used relatively
accurately for predicting the difference CBR values from the136Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Compressibility Characteristics of Residual Soil Subjected to Microbial
Treatment
Lee Min Lee1, Ng Wei Soon1, and Yasuo Tanaka1
Faculty of Engineering and Science, Universiti Tunku Abdul Rahman, Malaysia

1

ABSTRACTMicrobially induced calcite precipitation (MICP) has recently emerged as a new sustainable technique for soil
improvement. The technique utilizes bio-activity to generate cement compounds into soils, and hence to improve theirinherent engineering properties. This paper investigates the effectiveness of the MICP technique in improving thecompressibility characteristics of a typical tropical residual soil (sandy silt). A species of Bacillus group bacteria,
namely B. megaterium was used to trigger the calcite precipitation. The soil specimens were treated for three durations,i.e. 24, 48, and 72 hours. The experimental results showed that the recompression index (C r ), preconsolidation stress(p c ), and total consolidation settlement (S c ) reduced with the increased amount of calcite precipitated / MICP treatment
duration. The total consolidation settlements (S c ) of the MICP treated soils were reduced by about 2-21%. The MICP
treatment, however, has minimal effects on the compression index (C c ) and the coefficient of compressibility (C v )under a high applied stress. This was probably caused by the breaking of the calcite inter-particle bonds when the soil
subjected to high stresses.
Keywords: Consolidation; microbial treatment; residual soil; calcite precipitation

1. INTRODUCTION
Microbially induced calcite precipitation (MICP) has recently
emerged as a new sustainable technique for soil
improvement. The calcites precipitated from bio-activity areused for cementing and clogging soil particles, and hence toimprove the overall engineering performances of the soils.
This technique has also shown promising applications inother construction materials, i.e. improvements of concrete
strength, concrete durability [1], [2], and brick durability [3].
Despite being a relatively young technology, studies
pertaining to the soil improvement through MICP had beenreported by numerous researchers. DeJong et al. [4] provided
an overview of potential applications of this new technique in
improving the engineering properties of soil. Various
microscopy techniques were used to quantitatively assess the
distribution of calcite during the soil improvement. Qian et al.
[5] used three types of bacteria to consolidate and cement
sand grains. They found that the precipitation program wasessential for obtaining an effective cementation in the sandbody. Qabany et al. [6] investigated the potential factors
affecting the MICP. They concluded that high chemical
efficiency of MICP can be achieved by supplying a low input
rate (less than 0.042 mole/liter/hr) of urea and calcium
chloride. Recently, Martinez et al. [7] attempted to upscale
the bio-mediated treatment process for in situ
implementations. They concluded that continued researchmust address several challenges associated with the field
implementations such as soil and pore fluid interactions,

Corresponding author: Lee Min Lee
Email: mllee@utar.edu.my, Contact: +603-41079802

bioaugmentation versus biostimulation of microbialcommunities, controlled distribution of mediated calcite
precipitation, and permanence of the cementation.
Most of the current studies on MICP used microbiallycatalyzed hydrolysis of urea to produce calcite [8]. The
bacteria suitable for this type of biological processes are ofurease positive type i.e. genera Bacillus, Sporosarcina,
Spoloactobacilus, Clostridium and Desulfotomaculum [9].
The urease enzyme produced from the urease positivebacteria is responsible for decomposing urea (CO(NH 2 ) 2 )through the hydrolysis of urea:
CO (NH 2 ) 2 + 2H 2 O  2NH 4 + + CO 3 2(1)The release of ammonium (NH 4 +) increases pH, andeventually creates an ideal environment for calcite
precipitation with the presence of calcium ion (Ca2+) from the
supplied calcium chloride:
Ca2+ + CO 3 2-  CaCO 3(2)
The calcite (CaCO 3 ) precipitated is responsible for improving
the inherent properties of soil through biocementation andbioclogging. Biocementation is defined as an improvement ofsoil strength by the production of particle-binding materials
through microbial means, while bioclogging is the reductionof hydraulic conductivity of soil or porous rock bypore-filling materials generated by microbial processes [10].
From the foregoing, it can be concluded that most of theprevious studies had focused on the biocementation andbioclogging of sand grains [4] - [6], [8]. Attempts on othertypes of soils i.e. silty or clayey soils are still very limited.
Besides, the efficiencies of MICP soil improvements wereusually quantified by the measurements of calcite precipitated,shear wave velocity, or direct measurements of unconfinedcompressive strength and permeability. The effect of MICP
on other engineering properties of soil such as

137

compressibility characteristics had never been reported in theprevious studies. To address these research gaps, this paper
attempts to investigate the compressibility behaviours of a
MICP treated silty residual soil through one-dimensional
consolidation tests.
2. MATERIALS & METHODS
2.1 Type of Bacteria
The urease-producing bacterium used in this study was B.
megaterium ATCC 14581. The bacterium has a diameter ofabout 2 - 5 μm.
2.2 Cementation Reagents
Cementation reagents serve as important ingredients forpromoting calcite precipitation. As shown in the equations (1)and (2), the ammonium (NH 4 +) and calcium (Ca2+) ions aredecomposed from urea (CO (NH 2 ) 2 ) and calcium chloride(CaCl2 ), respectively. It is thus important to supply sufficientamounts of urea and calcium chloride into the soil to promote
effective MICP. In the present study, the concentration of
cementation reagents was fixed at 0.5 M [6]. The cementation
reagents also contained 3 g nutrient broth, 10 g NH 4 Cl, and2.12 g NaHCO 3 per litre of deionized water [11].
2.3 Soil Materials
The soil specimens used in the present study was a typicalsilty residual soil retrieved from a site in Kuala Lumpur,
Malaysia. The soil has particle sizes ranging from clay
fraction to 2 mm. Rebata-Landa [12] suggested that the most
optimum range of grain size for the bio-cementation process
is between 50 and 400 µm as bacterial activity cannot takeplace in very fine soils, while large amounts of calcites are
required to promote effective improvements in the very
coarse soil. In the present study, 32% of the residual soil
particles fall within the range of 50-400 µm. Table 1 tabulates
the physical properties of the soil.
2.4 Apparatus Setup
Fig. 1a shows the schematic diagram of the apparatus setup
for the MICP treatments. The apparatus consisted of a steel
mould of 50 mm in diameter and 170 mm in length (Fig. 1b),
an air compressor, a pressure tank, and an effluent collector.
The air compressor and pressure tank was used to control theflow pressure of the cementation reagents at approximately 1bar.
2.5 MICP Treatment Procedures
To perform the MICP soil improvement, the bacteria werefirst cultivated in nutrient broth to a concentration of 1×108
cfu/ml (optical density of 3.3). The bacteria were then mixed
with soil specimens and compacted into the prefabricatedsteel mould at a dry density of about 1519 kg/m3 (90% of
maximum dry density). The treatment commenced bysupplying the cementation reagents into the soil specimen atroom temperatures (ranging from 25 to 30o). Upon thecompletion of the treatment, the soil was extruded from thesteel mould, and was fitted into the standard oedometer ringof same diameter. The consolidation test was then carried out
in accordance to the British Standard (BS1377-Part 5).

2.6 Experimental Variables
The main variable in this study was treatment duration. Threetreatment durations were adopted, i.e. 24, 48, and 72 hours. It
was hypothesized that longer treatment duration would
promote more calcite precipitations, and thus thecompressibility characteristics of the soil for different
amounts of calcite can be assessed.
(a)Reagent
Pressureflow
tankPressurized
Valve
air
Pressure
Fabricatedregulator
steel mould
Air
Reagent
compressor
Effluent(b)

Influent

50 mm 170 mm
Effluent
Fig. 1(a) Schematic diagram of the apparatus setup;(b) Photograph of the fabricated mould
Table 1 Original Soil properties
Properties

IndicesGrain size

0% Gravel, 38% Sand,
43% Silt, 19% Clay
40.4%
24.9%

Liquid limit
Plastic limit
Maximum dry
density1688.5 kg/m3

3. RESULTS AND DISCUSSIONS
3.1 Calcite contents
Fig. 2 shows the correlations between treatment duration andcalcite precipitated. Acid-treatment weight loss technique
was employed to determine the carbonate content in the soil
specimens. The carbonate content that existed naturally in theoriginal soil was 0.67%. The increment in the carbonate
contents upon the MICP treatments can be safely assumed tobe caused by the calcite (Calcium carbonate) precipitation.
The calcite content was found to increase proportionally withthe treatment duration up to 48 hours. Beyond this duration,
the amount of calcite precipitated was negligible. A plausible
explanation to this observation is that the pore throats might
be clogged by the calcite precipitated resulting in lesser
cementation reagents flowing through the soil column. The
cementation reagents also contained nutrients for the bacteria.

138

The lack of nutrient supplies may create an unfavorable
environment for the bacteria to sustain the calcite
precipitation.

Fig. 2 Correlation between treatment duration and amountof calcite precipitated
3.2 Compressibility Characteristics of Soils
The compressibility characteristics of the soils were
evaluated through the following four parameters: (1)compression index (C c ) and recompression index (C r ) that
explains the relation between applied stress and amount ofconsolidation, (2) preconsolidation stress (p c ) that explains
the consolidation status of soil, (3) total consolidation
settlement (S c ) that indicates the absolute value of settlement,and (4) coefficient of consolidation (C v ) which explains thetime rate of consolidation.
Fig. 3 shows the plotting of void ratio (e) versus stress inlogarithm scale (log p) for the residual soils subject to various
MICP treatment durations. The values of all thecompressibility characteristics are summarized in Table 2.
The results showed that there was no consistent relationshipbetween the C c values and the amounts of calcite precipitated/ treatment duration. The C c values were in the range of 0.175– 0.195, indicating that the effect of MICP treatment on theC c value was of minimal. On the contrary, the C r values weresignificantly reduced (from 0.130 to 0.076) by the MICP
treatments. The higher the amounts of calcite precipitated, thelower the C r value. Similar trends were observed for the p c
and S c values. These observations can be explained by thecalcite bonding between the soil particles. Upon the MICP
treatment, the stiffness of soil was enhanced through theformations of calcite bonds. As the result, the amount ofconsolidation was reduced when the soil was subjected to alow stress (lower than p c ), and hence yielded a lower C r value.
However, the calcite bonds were insufficiently strong to resista high stress (higher than p c ). Consequently, the breaking of
the calcite bonds caused the soil to behave like its originaluntreated state. This explained the minimal effect of the
MICP treatment on the C c value. The total consolidation
settlement, S c of the MICP treated soils were reduced byabout 2 – 21% compared to the original soil. The reductionsin the S c were mainly attributed to the reductions in the C rvalues.
To evaluate the coefficient of consolidation (C v ), thedeformation of soil (δ) was plotted against square root of time(√t) under a stress lower than p c (i.e. 25 kPa), and a stresshigher than p c (i.e. 400 kPa), as shown in Fig. 4a and 4b,
respectively. The coefficient of consolidation (C v ) of soil was

estimated using the Taylor’s square root of time method [13].
Under the stress condition of lower than p c , the dissipation
rate of excess pore-water pressure in the MICP treated soils
were faster that of the original soil. The coefficient ofconsolidation (C v ) increased with the increased amounts ofcalcite precipitated / treatment durations. This phenomenon
can be attributed to the higher stiffness of the MICP treatedsoils compared to the original soil [14]. It is worth noting thatthe permeability of the MICP treated soils might be reduced
through bioclogging, and eventually causes a reduction in thedissipation rate of the excess pore-water pressure. However,
it is believed that the significant increment in the soil stiffness
associated to the formations of calcite bonds hasoutperformed the factor of reduced permeability. Under the
stress condition of higher than p c , the improvements in the C v
values were negligible. These observations can again be
explained by the breaking of the calcite bonds under the highapplied stresses.

Fig. 3 Void ratio (e) versus stress in logarithm scale (log p)

139

Table 2 Compressibility characteristics of soils
Compressibility
characteristics
Compression
index, C c

Treatment Conditions
Original
24
48
72
hrs
hrs
hrs
0.187

0.195

0.175

0.182

Recompression
index, C r

0.130

0.098

0.079

0.076

Preconsolidationstress, p c (kN/m2)

72

52

42

42

3.22

3.15

2.60

2.55

Coefficient ofconsolidation, C v(mm2/min):@ 25 kPa

85

133

236

236

@ 400 kPa

85

105

94

94

Total
consolidation
settlement, S c
(mm)

(a)

5. ACKNOWLEDGMENT
This work is financially supported by the Ministry of Higher
Education, Malaysia under Fundamental Research Grant
Scheme, FRGS.
6. REFERENCES(b)

Fig. 4 Deformation versus square root of time under theapplied stresses of (a) 25 kPa, (b) 400 kPa
4. CONCLUSION
This paper presents experimental works to investigate theeffectiveness of MICP treatment in improving thecompressibility characteristics of a typical residual soil. The
following conclusions can be drawn from this study:i)
The calcite content increases proportionally with thetreatment duration up to 48 hours. Beyond this duration,
the amount of calcite precipitated is negligible. This canbe attributed to the lack of cementation reagents andnutrients flowing through the soil column due tobioclogging.
ii) The recompression index (C r ), preconsolidation stress(p c ), and total consolidation settlement (S c ) reduce with
the increased amounts of calcite precipitated / treatment
durations. The total consolidation settlements (S c ) of the
MICP treated soils are reduced by about 2-21%.
iii) Under a low applied stress (lower than p c ), thecoefficient of consolidation (C v ) increased significantlywith the increased amounts of calcite precipitated /treatment durations. This can be explained by theenhanced stiffness of the MICP treated soils.
iv) The effects of MICP treatments on the compression
index (C c ) and the coefficient of consolidation (C v )under a high applied stress are negligible. This isprobably caused by the breaking of the calcite bonds
under high stresses.
v) Future studies may be carried out by applying unloading
and reloading cycles to further investigate the effect of
cementation on the soil compressibility after large
loading. Comparisons may also be made between thecompressibility behaviours of the MICP treated soils
and the natural aged clay.

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[2] Achal V, Pan X & Özyurt N, “Improved strength anddurability of fly ash-amended concrete by microbial
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[5] Qian C, Pan Q & Wang R, “Cementation of sand grains
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[9] Kucharski ES, Cord-ruwisch R, Whiffin V & Al-thawadi
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[13] Taylor DW, Fundamentals of soil mechanics. New York:John Wiley & Sons Inc., 1948.
[14] Yin JH & Fang Z, “Discussion on Physical modeling of
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140

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Strength and aging of cement treated low plastic soils
Fabien Szymkiewicz1, Antoine Guimond-Barrett2, Alain Le Kouby1, Philippe Reiffsteck1, and Sonia Fanelli1
1
IFSTTAR / Université Paris-Est, France
2
IFSTTAR / Université du Havre, France

ABSTRACT
Deep mixing is a method frequently used for various applications ranging from ground improvement and confinement
to foundations. With the increase of the use of deep mixing for foundations, it is imperative to better define themechanical behaviour and ageing of the material, which depends on the cement content and grain size distribution. This
study aims to quantify the impact of these parameters as well as the percentage of fines on the mechanical behaviour ofthe mixed material. A number of unconfined compression tests were carried out. The results showed that it was possible
to predict the strength of the mixed material after seven and 28 days of curing, based on the cement content and thepercentage of fines.
Keywords: deep soil mixing, unconfined compression tests; mechanical behaviour; sand, cement

1. INTRODUCTION
Initially, the main purpose of deep soil mixing was to improve
the stability and reduce settlements of structures such asembankments on soft soils of low shear strength or very highmoisture contents [1]. These days, improving the strength anddeformation properties as well as the permeability of very soft
soils by deep soil mixing is a frequently used stabilization
method, and the interest in the use of this technique not only
for soft soils stabilization but also to construct temporary
foundation/structural (load bearing) elements [2] andexcavation retaining walls is increasing, as the execution iseasier and less costly than traditional methods.
Deep mixing is an all-purpose term for a large number oftechniques in which binding agents are mechanicallydispersed within the soil in slurry form. The amount of
cement used in soil stabilization is usually much lower than
the amounts used for the construction of structural elements.
Consequently, the cement content varies largely from one
application to another, ranging from 50 to 500 kg per cubic
meter of soil. However, no binder content methodology has
yet been accepted, even if a large number of studies using
cement as well as industrial by-products such as wastepaper
sludge ash, pulverised fuel ash, etc. have been carried out([3]; [4]; [5]).
Sandy soils, that simplify the distribution of cement [6], arethe most suitable for new applications of deep mixing, as thecreated material can be compared to a mortar, although with ahigher water content. However, grain size distribution
influences the unconfined compression strength as it seems
that the smallest 25% to 40% of particle size controls themechanical behaviour of soils ([7]; [8]; [9]).
Furthermore, even if laboratory tests are the usual tests

carried out in feasibility studies for soil mixing projects and if
a large number of factors (such as the binder and soil nature aswell as the mixing and curing conditions) are known to
influence the strength and deformation properties of thetreated soils [10], no international standard exists for the
preparation of treated soil specimens in the laboratory [11]. It
is therefore difficult to compare results from differences
sources of the literature.
This communication focuses on the mechanical properties of
non-plastic sandy soils stabilized with cement to create
self-compacting mixes simulating the deep mixing process.
The objectives of this research were to evaluate and compare
the effects of time, cement content and grain size distribution
on strength of sandy soils when mixed in the laboratory withportland blastfurnace cement. Correlations and mathematical
relations between different parameters such as strength,
cement and fine content are presented, enabling the final
design strength to be predicted, depending on the cement
content and the grain size distribution.
2. EXPERIMENTAL PROGRAM
2.1 Materials
The soils used in the testing were obtained from different
regions of France, depending on the availability of test sites:
Fontainebleau sand, which is a French reference sand, comesfrom the south of Paris. Triel sand comes from the west of
Paris and Fréjus sand comes from the south of France, wherein situ soil-mixing tests were carried out [2].
Fontainebleau sand is a sub-rounded silicate sand which has auniform grain size distribution with no fines. Silica flour is an
artificial soil created by crushing Fontainebleau sand, andwas used in this research to study the influence of high finecontent on the strength of the soil mixing material, by using it
pure and also mixing it with Fontainebleau sand. Triel and141

Fréjus sands, which are also silicate sands, show distributions
that are more widely spread and contain 3% and 10% of fines
respectively. Grain size distributions of all the soils are shownin Fig 1.
100

Fontainebleau

Percent Finer by Weight

Triel
80

Fréjus
Silica Flour

60

SF50 - SilicaF50
SF75 - SilicaF25

40

20

0
0,001

0,01

0,1
Grain Size (mm)1

10

diameter, and, to avoid air bubbles in the specimens, wererodded and tapped. References [14] and [15] showed thismoulding method gives the best homogeneity and continuity(and hence compression strength) to the samples. The moulds
were capped, and sealed in a hermetic bag containing a highrelative humidity. These were stored at a temperature of 20 ±3°C until their testing day (7, 14, 21, 28, 56 and 90 days).
2.3 Unconfined compression tests
The samples were taken out of the moulds the day of the test.
They were then cut and smoothed to create plan and parallel
end surfaces. They were measured and weighed, then tested
for their unconfined compressive strength (q u ). Three
specimens were tested at each curing time. The mean value ofthese tests is presented in this paper. In cases of clearly
deviating results, an additional specimen was tested.

Fig. 1: Grain size distribution of the soils
These soils present Methylen Blue Values ranging from 0(Fontainebleau sand) to 0.2 (Fréjus), showing the low content
of clay particles.
The cement used for this experimental program is a Portland
blastfurnace cement containing 85% ground granulated blast
furnace slag, with the rest Portland clinker and a little gypsum(European classification: CEM III/C 32,5 N CE PM-ES NF‘HRC’) [12].

3 RESULTS AND ANALYSES

2.2 Mixing, moulding and storage procedure
The cement content C was chosen to cover the complete
range of French soil mixing applications (from ground
improvement to structural elements), i.e. from 70 to 400
kg/m3. Cement contents and equivalences are presented inTable 1.
Table 1: Cement equivalences
C (kg/m3)
C (%)
70
4.2
140
8.4
210
12.7
265
16
320
19.4
400
24.2

qu = a × C b

It has been observed that as expected the cement content has agreat effect on the strength of these sand–cement mixtures. A
small difference in cement content has a significant impact on
the performance of the soil–cement mixing.
Reference [16] proposed a power function, defined by Eq. (1),as the most adapted to fit the experimental relation between
the unconfined compression strength (q u ) and the cement
content (C).
(1)

where a is a parameter expressed in kPa and b a dimensionlessparameter. Both are experimental parameters.
Fig. 2a and 2b show the relationships between q u and C for
Fontainebleau sand and the Silica Flour. Again, the best fit
curves follow a power function. However, the curvature is not
the same, depending on the type of sand and age. The results
show the same pattern for the Fréjus and Triel sands.
The power function is the best fit for the range of cement
contents tested (up to 25%, which is close to the maximum
cement content used for deep mixing applications). However,
it is clear that should C increase drastically, a plateau would
be reached, then a decrease of strength would appear,
meaning that the power function would no longer be the best
fit.

142

18
15

7 days old
14 days old

12qu (MPa)

The water content w was chosen to attain a sufficient
workability, i.e. to ensure that the material was fluid enoughto be poured in the moulds and be self-compacting.
The preparation and conservation method of the specimens is
described in a few words here. More details can be found in[13]. Soil and cement were first thoroughly dry-mixed
manually, in order to obtain a uniform consistency. Thispremix was then put in the Hobart mixer and water wasadded. The amount of water was calculated based on thetarget water content of the mixes, i.e. 19% for the pure sands.
After waiting three minutes to ensure complete wetting of thematerials, the mixing process could now begin: five minutes
in the mixer with constant intervention of the operator. The
mix was then poured into cylindrical moulds of 52 mm

21 days old

928 days old
90 days old

6

180 days old

3
0
0

5

10
15
Cement (%)

20

25

2,5

12

7 days10890 days
1,5

6

FréjusTriel
Fontainebleau
SF75-SilicaF25
SF50-SilicaF50
SilicaF

b7

qU (MPa)2

28 days

1

4
0,5

2
0

0

5

10
15
Cement (%)

20

25

0,1

1

10

100

C63 (%)

Fig. 2: Evolution of the qu–C relationship with time for (a)Fontainebleau sand and (b) silica flour
Whereas the mixtures made with pure Fontainebleau sand do
not show a noticeable change of the values of parameters aand b (with a increasing from 7 to 28 and b decreasing from
2.15 to 1.96 between 7 and 90 days of curing), comparatively,
the Triel and Fréjus mixtures show a significant variation for
those two parameters with time, with a increasing from 30 for
both sands to 1144 for Triel and to 210 for Fréjus, and with b
decreasing from 1.8 to 0.9 and from 1.5 to 1.17 for Triel andFréjus respectively. For the mixtures containing silica flour, aincreases significantly with time, whereas b is relatively
stable.
The decrease of b with time means that the effect of thecement content decreases for older ages.
A clear linear relation in a semi logarithmic plan between theparameters a and b exists after 7 days of curing (Fig. 3). Eq.
(2) describes the relationship between a and b after sevendays of curing.
(2)b7 = −0.3145 ln a 7 + 2.7097

Fig. 4: Relation between C 63 and b 7
It is therefore possible to propose a formula allowing thedesigner to estimate the strength of stabilised granular soils
based on the cement content and the percentage of fines (Eq.
(5)).

qu7 = 19.428 × C 1.777 −0.171 ln C63 × C 630.544(5)This formula gives compression strength after 7 days ofcuring. Fig. 5 and Eq. (6) show the good correlation between
the laboratory results and the values given by Eq. (5).
10predicted qU7 (MPa)0FréjusTriel
Fontainebleau
SF75-SilicaF 25
SF50-SilicaF50
SilicaF

8
6
4
2
0
0

2

4
6
measured qU7 (MPa)810Fig. 5: Predicted q u7 versus measured q u7

2,5

qu 7 predicted = 0.96 × qu 7 measured

2
Fréjus
1,5

b7

Triel
Fontainebleau

1SF75-SilicaF25
SF50-SilicaF50

0,5SilicaF
0
1

10

100

1000

a7

Fig. 3: Parameter b 7 as a function of parameter a 7
Therefore, it is possible to produce a formula (Eq. (3)) linking
q u to C with only b 7 as an experimental parameter:

qu = e 8.616 × e b7 (ln C −3.18)(3)Introducing the fines content C 63 (63 μm is a common
threshold extensively used in earthworks to separate finefrom coarse particles [17], it is possible to find a relation
between b 7 and the percentage of fines (Eq. (4)).
(4)
b7 = −0.1711 × ln C 63 + 1.7767
Fig. 4 shows the relationship between b 7 and the percentage
of fines C 63 .

(6)

We have established that it is possible to predict the strength
after 7 days of curing for a non-plastic granular soil,
depending on the content of cement and fines. This isvaluable as contractors have a need to determine the strengthof the material as quickly as possible (so as to be able toreadjust the cement content for example). However, fordesign purposes, it is often the unconfined compression
strength at 28 days that is taken into account; therefore it isimportant to be able to predict this strength.
Reference [18] showed that q u28 is by no means the highest
achievable strength, and that it depends from the grain size
distribution of the soil. Therefore, considering q u28 equal to 2
times q u7 for purpose design as estimated by [6] can be
dangerous. In our case, doing so leads to an overestimation ofq u28 by almost 10%. On the contrary, it was shown by [18]that the highest achievable strength is at least equal to 2 times
q u7 , for treated sandy soils.
Looking closer to the shape of the typical curves presenting
the strength as a function of time, we observe that the relation
linking these parameters follows a logarithmic function (Eq.
7)

143

qu t = g × ln(t ) − h(7)Where g and h (in kPa) are experimental parameters.
Reference [19] showed that the values of these parameters arein general close to each other, and directly related to q u7 . We
can then propose the following formula (Eq. 8)(8)
qu t = β × qu 7 × (ln(t ) − 1)
With β an experimental coefficient to be determined, andimplicitly taking into account the water and cement content.
During the curing process, the only known variable value isthe water content, which decreases. Reference [20]proposed a formula enabling us to estimate the final watercontent of the mix (Eq. 9).

wn
C
−k×
wn + 1
1000
wf =
1
C
+ (1 + k ) ×ρ soil ×
wn + 11000

ρ soil ×(9)

Where w n is the initial water content and k the non evaporable
water content, equal to 0.23. Considering that the parameter β
is equal to the ratio final water content wf on initial watercontent wi , we have the following formula (Eq. 10).

qu t =

wf
wi

× qu 7 × (ln(t ) − 1)

(10)Finally, Fig. 6 shows the comparison between the strength
measured after 28 days of curing and the strength at 28 days
calculated from the Eq. 11, which combines Eq. 5 and Eq. 10:
the precision is very good (R² = 0.88) with an underestimationof the laboratory results of about 6 %.

q ut =

wf

× 19.428 × C 1.777 −0.171ln C63 × C 630.544 × (ln(t ) − 1)wi

qU28 predicted from the predicted
qU7 (MPa)

16

(11)

FréjusTriel
Fontainebleau
SF75-SilicaF 25
SF50-SilicaF50
SilicaF

12

4

0
4

5 ACKNOWLEDGMENT
The authors would like to thank Soletanche-Bachy forfunding this research.
6 REFERENCES

8

0

strength of the cement mixed materials.
This study has shown that the mix strength depends on the
cement content, following a power relationship. In addition,
the exponent depends directly on the soil fines content: thecleaner the sand, the higher the value. It seems that the upper
limit of the value of this parameter is around 2.1 and the lower
limit is 0.8.
By studying six granular soils, we have proposed a formula topredict a cement treated granular soil q u7 solely from thecement content and the parameter C 63 .
Finally, a formula to predict 28-day strength is proposed,
based on strength after seven days of curing, with verysatisfactory accuracy. We are therefore able to predict 28-day
strength of a granular soil treated with cement knowing only
soil particles size and the cement content used. The results
were not compared to other authors’ results due to differencesin the preparation method. In particular, compaction is a big
issue. However, the relationship linking q u to C is always apower function for the range of applications concerned.
Therefore, the formula simply needs to be reviewed forcompacted Soil Mixing applications, taking into account thecompaction energy.
This formula needs now to be tested in situ. However, it
should not be a big issue to transpose it for on sites mixes, as
the sandy soils are the easiest soils to mixes, with a smaller
number of inclusions [18] and a ratio in situ / laboratory witha value near one [20]. Hence, this is a real advance forpractitioners, when encountering non to low plastic soils. It is
now important to study the influence of plastic fines on
strength and aging.

8
qU28 measured (MPa)1216Fig. 6: q u28 predicted with Eq. (5) and (10) versus q u28
measured in the laboratory
It is then possible to predict the strength of the Deep Mixing
material made of a non-plastic sandy soil and cement,
knowing only the fine content C 63 of the soil, the target
cement content C (%), and the initial and final water contents
wi and wf of the Deep Mixing material.
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Research Center [in Japanese], 1999.

145

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Performance Of Cantilever Sheet Pile Supporting Excavation Adjacent To
Strip Footing
Yousry M. Mowafe 1, Ahmed M. Eltohamy 2 and Ahmed A. Mohamed 3
1 Professor of Geotechnical Eng., Faculty of Engineering, Azhar Univ., Egypt.
2 Associate Professor of Geotechnical Eng., Faculty of Engineering, Beni Suafe Univ., Egypt.
3 Assistant Professor of Geotechnical Eng., Faculty of Engineering, Azhar Univ., Egypt.
ABSTRACT
Ground movements induced from excavations are a major cause of deformation and damage to the adjacent
foundations and utilities. A numerical prototype model has been verified by analyzing the results of an experimental
model. The influence of foundation depth and distance from the excavation on straining actions and deformation of
sheet pile and strip foundation were investigated. The lateral deformation of the cantilever retaining sheet pile walland the accompanied vertical settlement of adjacent foundation nearly vanished as the excavation was shifted far
away from the foundation by a distance equal to or more than 1.5 times excavation depth in both medium and dense
sand. As the ratio between adjacent foundation depth to excavation depth increases a corresponding decrease instraining actions on retaining sheet pile and settlement of foundation occurs. Differential settlement of strip
foundation adjacent to excavation vanished when distance of the footing reached 1.5 times excavation depth.
Empirical formulae have been presented to express head deflection of cantilever sheet pile retaining wall, maximum
bending moment and ratio of adjacent strip footing maximum settlement to corresponding values in the case of no
adjacent footing or in case of footing before excavation cases.
Keywords: Braced excavation, Sheet pile, Lateral deformation, Straining actions.
1. INTRODUCTIONSpace of construction activities is usually limited due to theproximity of adjacent structures. A major concern for projects
involving deep excavation is the impact of such excavation on
adjacent ground and structures [1]. Deformations of the
excavation support system, adjacent ground and structures are
influenced by a number of factors including stiffness of the
excavation support system, installation procedures of thesystem, soil conditions, excavation procedures, type offoundation, distance of foundation from excavation, andorientation of foundation with respect to the excavation [2-4].
Wall pressures and deformation increase with the decrease inadhesion between sheet pile wall and soil [5]. The
deformation behind flexible retaining walls anchored at the
top was studied analytically through performing a number ofmodel tests with dense sand [6]. Using a series of numericalnonlinear finite element analyses of experiments the effectsof wall embedment depth, support conditions, and stresshistory profile on the undrained deformations around a braced
diaphragm wall in a deep clay deposit were studied [7]. The
results were presented in the form of design charts for
estimating ground movements as functions of the excavation
depth, supporting conditions. Braced excavation wasobserved to inevitably result in deformation of the adjacent
ground and settlement of adjacent buildings behind the wall.
Building settlement is believed to exceed the amount ofassociated ground settlement by 500% of the settlement underbuilding weight before excavation [8]. This is due to thedecrease in stiffness of the foundation soil that is caused bylateral movement of the excavation face. It was also observedthat additional settlement substantially decreased as thefoundation level reached a depth equal to or more than 0.5 of
excavation depth. To quantify the amount of building
settlement, a model testing was conducted to determine themagnitude of the additional settlement of the structure and theeffect of varying the building’s depth and of its distance awayfrom the excavation wall. Bearing walls oriented in a
direction perpendicular to an excavation wall tend to become
distorted with shear strain and lateral strain at the foundationlevel [9]. When ground movement initially impinged on thefront of the building, the building is primarily subjected toshear distortion and lateral strain at the base. For deep
excavations in stiff clays, residual soils, and sands, theaverage normalized lateral wall movement divided by
excavation height “ζ Hmax /H” nearly equaled 0.2%, andincreases to 0.5% for poor construction conditions [10-12].
Other researchers specified a range of “0.1-0.5%” for“ζ Hmax /H”, [13-14]. “ζ Hmax /H” is affected by relative densityof soil. For loose and dense sand, it was observed to range
from “0.2% - 0.4%” and “0.1% - 0.2%”, respectively, [15].
Adjacent foundation settlement was observed to depend on
footing load level and relative density of sand. Excavation
effect nearly vanished as the foundation was shifted away
three times the foundation width [16]. There are two ways ofmodeling the interaction between the soil and the structure
including a method based on a modified Winkler model inwhich the soil reaction is expressed by a number ofelasto-plastic supports. The other method models the soil
reaction as continuous elato-plastic media. The first method

146

underestimates forces in struts and anchors, as well as
bending moments in walls [17].
The present study aims to investigate different aspects of the
behavior of excavations in soil and the response of structures
located near these excavations. The main objectives for thisresearch are:I. Establishing an analytical model to represent realistic
behavior for the excavation and the adjacent strip footing.
II. Studying the parameters affecting the behavior of thechosen model such as distance between the foundationand excavation and foundation depth.
III. Expressing maximum deformation and straining actions ofsheet pile and settlement of foundation adjacent toretained excavation in terms of corresponding values ofbefore excavation.
2. MODEL TESTPROCEDURE

SETUP,

MATERIALS• Depth of excavation = (H) cm,• Total height of sheet pile wall = (d) cm,• Building distance from excavation = (x) cm,• Foundation depth= (D f ) cm.
R

R

AND

A schematic drawing of the test setup used in this study isshown in Fig. 1. The model test box was made of 2 mm thicksteel sheets, 500 mm wide, 1000 mm high, and 1700 mm
long. The excavation wall was made of a 1 mm thick steel
sheet. The model strip footing was made of 500 mm firm
wooden block. An automatic IOtech DaqBoard/2005 dataacquisition system with LabView software was used tocontinuously monitor and record data during all stages of thetests. Multiplexing Isolated Analog Input DBK207 with 16
channel capacity was used to monitor the output voltages
from instruments.

Fig. 2. Sheet pile wall supporting excavation and Data
Acquisition system.
The undrained shearing resistance of the sand compacted to a
dry unit weight of 17 kN/m3 was determined using triaxial
tests and found to be 36°. The gradation curve is shown in
Fig. 3. According to the Unified soil classification system,
sand can be classified as SP. More detailed information
regarding the experimental setup may be found in reference
No. [18].
P1000 mmP

3. NUMERICAL MODELING AND SELECTION OF
PARAMETERS
Soil code Finite Element Program Plaxis 8.0 [19] was used innumerical modeling. In order to determine the appropriate
numerical modeling meshing coarseness and refinement thatare most suitable in modeling the studied geotechnicalapplication a number of program runs were carried out toanalyze the experimental model studied cases with different
mesh configuration. The configuration that results in similar
performance and close values of dependent parameter uponvariation of independent investigated parameters will be1700 mmFig. 1 Model of excavation, adjacent area and footing.
The data acquisition system was connected to a PC Pentium
IV processor, as shown in Fig. 2. The footing was stiffened bya vertical tapered wooden plate 500 mm wide extending fromstrip footing top level to the level arm loading rod. The
dimensions of different model items are as follows:• Width of excavation (w) = 16 cm,

147

adopted for extension of the study with prototype dimensions.

The interfaces allow for the specification of a reduced
wall friction compared to the friction in the soil.
Table 2. Material properties of sheet pile.
Parameter
Normal stiffness (EA) (kN/m)
Flexural rigidity (EI) (kN/m2/m)
Equivalent thickness (d) (m)Weight (w) (kN/m)Poisson’s ratio (υ)

Value
2.68 x106
9.68x103
0.208
1.2
0.3

3.3 Strip Footing
Strip footing can be simulated as a beam element taking themodulus of elasticity for concrete “E” as follows:
Fig. 3. Particle size distribution of tested soil.

E = 4400

Fcu (N/mm2)

3.1 Material Model of Sandy Soil

Where

In order to make realistic predictions of the stability anddeformations of excavation and adjacent strip footing, thehardening soil model in PLAXIS program was used for sandidealization. Material properties are presented in Table 1 formedium and dense sand. The sand is modeled by 15-nodetriangular elements in the analysis using an elastic perfectly
plastic hardening model. The value of relative density is 0.53
for medium sand and 0.76 for dense sand.

F cu : characteristic strength (25N/mm2)

Table 1. Input Physical And Mechanical Properties of Sand.
Parameter
Medium Dense
sand
sand
Soil dry unit weight (γ d ) (kN/m3)
17
18
Soil wet unit weight (γ wet ) (kN/m3)
20
20
Secant stiffness in drained Triaxial test
14925
29850
(Eref 50 ) (kN/m2)
Tangent stiffness for odometer loading
10447
20894
(Eref oed ) (kN/m2)
Unloading / reloading stiffness (Eref ur )
44775
89550
(kN/m2)
Poisson’s ratio for unloading-reloading
0.25
0.25
(υ ur )
Cohesion (c) (kN/m2)
1010Friction angle (φ) (degree)36
42
Dilatancy angle (ψ) (degree)
612Interface reduction factor (R)0.8
0.8Power for stress-level dependency of
0.5
0.5
stiffness (m)Reference stress for stiffness (pref)
100
100(kN/m2)
3.2 Sheet Pile Wall
The sheet pile wall “S.P.W.” is modeled as a beam
element. The parameters for sheet pile wall are presented
in Table 2. The interaction between the wall and the soil
is modeled at both sides by means of interface elements.

(1)

The footing is loaded with distributed load 100 kN/m2,
the properties of footing are given in Table 3.
Table 3. Material properties for strip footing.
Parameter
Value
Axial stiffness (EA) ( kN/m)
3.811E+07
Flexural rigidity (EI) (kN.m2/m)
3.175E+06
Equivalent thickness (EI) (m)1.00
Weight ( w) ( kN/m)
25.00
Poisson’s ratio ( υ)
0.2
3.4 Verification of Numerical Modeling
The mesh coarseness and refinement that results in similar
performance of experimental and numerical small-scalemodel analysis is the very fine mesh coarseness andrefinement along “S.P.W.” line. Figures 4 and 5 illustrateexperimental and numerical bending moment “B.M.”
distribution along “S.P.W.” for Rx = (x/H) equals 0.5 and 1,
respectively, for medium sand. As can be observed thegeneral trend of “B.M.” distribution along “S.P.W.” for bothexperimental and numerical analysis was similar. Maximum“B.M.” of experimental and numerical curves occurs at one
third “S.P.W.” height measured from its top level. The
maximum experimental “B.M.” values were observed to behigher than corresponding numerical values by nearly 16%.
The previously presented observations regarding “B.M.”
distribution and maximum “B.M.” value applies when the
strip footing was shifted away from the “S.P.W.” with thecases of Rx = 1.5 and 2.0. The reduction of maximum “B.M.”
value due to the increase of Rx compared to the closest caseof Rx = 0.5 are 68, 90 and 94% for Rx = 1.0, 1.5 and 2.0,
respectively, for the experimental investigation . The
corresponding values recorded for the numericalinvestigation are 65, 90 and 93%.

148

presented in terms of “R x = 0.5, 1, 1.5 and 2” and R d having
the values of “0.25 and 0.5”.
R

R

R

R

8

7

6

5

4

3

2

Case No.

1

Table 4. Cases analyzed in the numerical study.

medium

1.50

2.00

0.50

1.00

11

12

13

14

R

2.00

1.00

Case No.

R

16

Rx = Χ Η

10

0.50
1.50

0.25

Η
0.50

R

9

R

15

Rd = Df

Dense

2.00

1.50

1.00

R

0.50
0.50

Rx = Χ Η
R

0.25

Η

2.00

R

1.50

R

1.00

Rd = Df

0.50

Fig. 4. Variation of the sheet pile wall bending moment for x=
8cm and H=16cm.

Fig. 5. Variation of the sheet pile wall bending moment for x=
16cm and H=16cm.
Comparing these ratios, it can be concluded that variation of“B.M.” values corresponding to variation of the independent
variable Rx recorded close values in both experimental andnumerical analysis. Hence it can be concluded that the
adopted mesh coarseness and refinement results in a similar“B.M.” distribution profiles and relatively close “B.M.”
maximum values. The variation in maximum “B.M.” values
resulting due to corresponding variation of the independent
parameter Rx has close values in both experimental andnumerical analysis. Hence the mesh configuration will be
adopted for the prototype scale model investigation of thetested parameters.
4.
NUMERICALINVESTIGATIONPROTOTYPE DIMENSIONSWITHThe general layout of the physical model is shown in Fig. 6.
The adopted finite element mesh was illustrated in Fig. 7. The
independent parameters taken into consideration are thedistance of strip footing “x” and the depth of foundation “D f ”.
Table 4 presents the cases of studied parameters including
relative depth “Rd= D f /H” and the ratio of strip footing
distance to depth of excavation “Rx=X/H” . The resulting
straining actions and deformations due to excavation are
R

R

R

R

R

Fig. 6. Model of excavation, adjacent area and strip footing.
5. ANALYSIS AND RESULTS
5.1 Lateral Deformation of Sheet Pille
Figure 8 illustrates the lateral deformat ion profile of “S.P.W.”
due to retained excavat ion in medium sand and the adjacentstrip footing loading wit h varying distances fro m “S.P.W.” of
Rx = 0.5, 1.0, 1.5 and 2.0, and Rd = 0.25. As may be expected
the maximum lateral deformat ion “ζ Hmax ” was recorded at
The head of the wall lateral deformat ion was sharply reducedwith depth until the mid height level o f “S.P.W.”. Lateral
deformation at this level reaches a value less than 50.0 % ofdeformation at head level wit hout an adjacent strip footingcase. The reduction ratio reaches a value of nearly 20% for Rx
values of 2.0, 1.5 and 1.0, while it reaches 10% for Rx = 0.5.
Lateral deformat ion gradually reduces from mid height level
unt il approaching nearly zero value at tip level o f “S.P.W.”.

R

149

R

R

Fig. 7. Finite element mesh for the model of deep excavation
and adjacent area.

Fig. 9. Variation of the lateral deformation of sheet pile wallfor different values of R x and “R d = 0.50” for medium sand.
R

Fig. 8. Variation of the lateral deformation of sheet pile wallfor different values of R x and “R d = 0.25” for medium sand.
R

R

R

R

Figure 9 illustrates the lateral deformation profile for
excavat ion in medium sand for different Rx values with Rd=
0.5. Figures 10 and 11 illustrate the same profile for
excavat ion in dense sand for Rd= 0.25 and 0.5, respect ively.
As can be concluded fro m these figures the lateral
deformation profile remains unchanged fro m that previously
described. Figure 12 illustrates the head deformat ion “ζ Hmax ”
values for the previously ment ioned different cases. It can beconcluded that the values of “ζ Hmax ” of “S.P.W.” excavated indense sand are nearly 50% of the corresponding values
recorded in case of medium sand. Other researchers observed
a reduction in head deformat ion of “S.P.W.” excavated indense sand equaling one half o f the corresponding value when
excavat ing in loose sand [15]. It can be also observed that theexistence o f strip footing in case o f close excavat ion of Rx=
0.5 produces increase in head deflection o f 2.4 and 3.0 times
the non strip footing case for excavat ion in medium and densesand, respectively. As the footing was shifted away fro m the“S.P.W.”, a reduction occurs in head deflect ion by
R

R

R

R

R

R

R

Fig. 10. Variation of the lateral deformation of sheet pile wallfor different values of R x and “R d = 0.25” for dense sand.
R

R

R

R

50, 40 and 30% for both medium and dense sand fo r Rx= 1.0,
1.5 and 2.0, respectively. The reduction in head deflect ion“ζ Hmax ” that occurs due to increasing foundat ion relat ive
depth of the footing fro m Rd of 0.25 to 0.5 for both mediu mand dense sand is 38, 25 and 20% for Rx of 0.5, 1.0 and 1.5,
respectively. This may be attributed to the following. As the
strip footing gets deeper the increase in vert ical stress and theaccompanying increase in lateral horizontal stress is shifted
downwards and the rotation is reduced result ing in less head
deflect ion. The effect of foundat ion depth vanis hes as thefooting is located relat ively far away fro m the “S.P.W.” with
Rx reaching 1.5. As for “S.P.W.” head deflect ion to
excavat ion depth ratio “ζ Hmax /H” the recorded values fordifferent Rx and Rd values as illustrated in Table 5. As can beobserved fro m Table 5, “ζ Hmax /H” values coinside with values
cited in the previous work [15], for Rx≥1.5. For strip footing
loading closer to “S.P.W.” (Rx= 0.5 and 1.0) the recorded
values of “ζ Hmax /H” were higher than the corresponding
values recorded in previous work [15].

150

R

R

R

R

R

R

R

R

(ζH max/ H ) ratio =For dense sand:(ζH max/ H ) ratio =(ζH max/ H ) S .F .
= (1.6 −1.22 Rd ) Rx (1.2 Rd − 1.2)
(ζH max/ H )(2)(ζH max/ H ) S .F .
= (2.0 −1.47 Rd ) Rx (1.2 Rd − 1.2)
(ζH max/ H )
(3)

Where

(ζH max/ H ) S.F. : Head deflect ion of “S.P.W.” in case ofan adjacent strip footing.
(ζH max/ H ) : Head deflection of “S.P.W.” in case ofwithout an adjacent strip footing.
5.2 Bending Moment along Sheet Pile

Fig. 11. Variation of the lateral deformation of sheet pile wallfor different values of R x “R d = 0.50” for dense sand.

Fig. 12. Variation of the lateral head deformation of sheet pilewall for different values of R x and Rd.
The effect of strip footing loading adjacent to “S.P.W.”
excavat ion may be expressed in terms of “ζ Hmax /H” value inthe case of a strip footing loading adjacent to thecorresponding value without an adjacent strip footing “forsheet pile embedment depth to excavat ion depth ratio
equaling 2.0” may be expressed as fo llows:
For medium sand:Table 5. Head deflection to excavation depth ratio “ζ Hmax /H”
for different Rx and Rd.
Medium Sand
Dense Sand
Rx
Rd
0.25
0.5
0.25
0.5
Non strip
0.4
0.375
0.163
0.165
footing
0.5
0.9
0.613
0.5
0.32
1.0
0.5
0.375
0.25
0.138
1.5
0.4
0.325
0.125
0.11
2.0
0.275
0.25
0.125
0.11

Figure 13 illustrates bending mo ment “B.M.” distribut ion
along “S.P.W.” due to retained excavation and adjacent strip
footing loading for medium sand with Rd= 0.25 and different
Rx values. As can be observed fro m this figure, “B.M.”
increases fro m head level recording a maximum value at 0.38
of “S.P.W.” height fro m top level. Then “B.M.” decreases
downwards unt il vanishing at 0.67 of “S.P.W.” heightmeasured fro m its top. This profile was observed also for thecase of Rd=0.5 for medium sand and Rd= 0.25 and 0.5 fordense sand. Figure 14 illustrates maximum “B.M.” values
with different Rx, Rd and sand densit y. As can be observed
fro m this figure, maximum “B.M.” value wit h the case of
excavat ion in medium sand was nearly 1.5 times thecorresponding value recorded with dense sand. It can be also
observed that placing a strip footing close to a retaining“S.P.W.” wit h Rx=0.5 results in increasing maximum “B.M.”
value by 2.6 and 2.8 for Rd=0.25 for excavat ion in mediu mand dense sand, respectively. Relat ively smaller increasing
ratios, of 2.0 and 1.8 were recorded for deeper foundat ion
level o f Rd= 0.5 due to shift ing of increase of lateral pressure
downwards. The matter results in increasing the fixat ion
effect of embedment length of “S.P.W.” and hence results in
reduction of “B.M.”. Shift ing of strip footing away fro m“S.P.W.” results in reducing maximum “B.M.” values by
0.55, 0.45 and 0.35 for Rx= 1.0, 1.5 and 2.0, respectively for
excavat ion in medium sand. Corresponding values for
excavat ion in dense sand are 0.6, 0.57 and 0.51, respect ively.
It can also be observed that as the strip footing is shifted 2.0
times excavat ion depth fro m excavat ion the footing existence
does not result in addit io nal straining act ions alo ng sheet pile
compared to the non strip footing case. The ratio between
maximum “B.M.” values in the case of an adjacent strip
footing to those in the case of no strip footing “for sheet pileembedment depth to excavation depth ratio equaling 2.0”
may be expressed as fo llows:
For medium sand:Max.( B.M .) ratio =Max( B.M .) S .F.
= (1.76 − 1.0 Rd ) Rx( Rd −1.0)
Max( B.M )

(4)
151

Fig. (15): Variation of raft settlement for different values ofR x and "R d = 0.25" for medium sand.

Fig. 13. Variation of the sheet pile wall bending moment fordifferent values of R x and “R d = 0.25” for medium sand.

Table 6. Differential Settlement for different Rx and Rd
values.
Rx
0.5
1.0
1.5
2.0
MediumDense

Fig. 14. Variation of maximum bending moment values on
sheet pile wall for different values of R x and Rd.
For dense sand:Max. ( B.M .) ratio =Max( B.M .) S .F .
= (1.88 − 1.25Rd ) Rx( Rd −1.0)
Max( B.M )(5)
Where:
Max. (B.M.) S.F : Maximum “B.M.” value with an adjacentstrip footing.
Max. (B.M.): Maximum “B.M.” value without an adjacentstrip footing.
5.3 Settlement of Strip Footing
Figure 15 illustrates variation o f strip footing settlement fordifferent Rx and "Rd=0.25" for medium sand. It can beobserved that the footing settlement increases as the footing iscloser to the retaining “S.P.W.”. The settleme nt of strip
footing towards excavat ion side was recorded to be 1.6 and1.2 times corresponding settlement in the case of no adjacentexcavat ion for Rx=0.5 and 1.0, respectively. Longer distance

Rd

0.25
0.50
0.25
0.50

4.67*10-3
4.0*10-3
2.71*10-3
1.33*10-3
P

P

P

P

2.0*10-3
2.0*10-3
0.67*10-3
0.8*10-3
P

P

P

P

0.0
0.0
0.0
0.0

0.0
0.0
0.0
0.0

was between strip footing and excavat ion ( cases of Rx=1.5
and 2.0), the settlement value remains nearly unchanged
compared to the corresponding value before excavation.
Table 6 illustrates recorded values of differential settlementfor different Rx and Rd values. As can be observed from thistable different ial settlement of strip footing exceeding 1/500
was recorded in the case of medium sand for Rx = 0.5 and 1.0
for both foundat ion level values of Rd= 0.25 and 0.5. This
was also recorded in the case of dense sand for relat ively
close footing having Rx=0.5 and a relat ively shallow
foundat ion level of Rd=0.25. No different ial settlement wasrecorded for strip footing having Rx=1.0 and 1.5 for bothstudied foundat ion levels and sand densit ies. For different Rx
and Rd values, it can be observed that the maximu msettlement value was nearly reduced to one half it s value wit h
the case of excavation in dense sand compared to the medium
sand case. It also can be observed that executing retained
excavat ion near the strip footing result s in increasing its
maximum settlement value by nearly 60% and 30% for Rd =0.25 and 0.5, respectively. Maximum settlement value
reduces as the footing was shifted away fro m the “S.P.W.”
unt il nearly reaching the value corresponding to that before
excavat ion case as Rx reaches 2.0. The ratio of maximu msettlement value in the case o f an adjacent excavation of avarying distance fro m strip footing Rx to the corresponding
value wit hout excavat ion (Max. (S) ratio) for sheet pileembedment depth to excavation depth ratio equaling 2.0”
may be expressed as fo llows:
For medium sand:152

Max. ( S ) ratio =(6)

For dense sand:Max. ( S ) ratio =[5]

Max( S ) S .F .
= (1.37 − 0.4 Rd ) Rx ( 0.5 Rd − 0.5)
Max( S )Max( S ) S .F .
= (1.29 − 0.3Rd ) Rx ( 0.5 Rd − 0.5)
Max( S )[7](7)
Where:
Max. (S) S.F: Maximum settlement value in case of an
adjacent strip footing.
Max. (S): Maximum settlement value in case o f without anadjacent strip footing.
6.

[6]

[8]

[9]

CONCLUSIONSIn the present numerical analysis study the effect of retained
excavation on adjacent strip foundation was investigated. The
effect of strip footing relative distance and foundation depth
related to cantilever sheet pile wall retained excavation on
straining actions along sheet pile and deformation offoundation was studied. Analyzing the investigation results it
may be concluded that the lateral deformation of thecantilever retaining sheet pile wall and the accompanying
vertical settlement of adjacent foundation nearly vanishes as
the excavation was shifted far away from the foundation by adistance equaling or more than 1.5 times excavation depth inboth medium and dense sand. As the ratio between adjacent
foundation depth to excavation depth increases a
corresponding decrease in straining actions on retaining sheet
pile and settlement of foundation occurs. Differential
settlement of strip foundation adjacent to excavation vanishes
when the distance between excavation and the footing
reaches 1.5 times excavation depth. Empirical formulae havebeen presented to express head deflection of cantilever sheet
pile retaining wall, maximum bending moment and adjacentstrip footing maximum settlement to corresponding values incase of no adjacent footing or before excavation cases for
sheet pile embedment depth to excavat ion depth ratio
equaling 2.0.

[10]

[11]

[12]

[13]
[14]

[15]
[16]

7. REFRENCES
[1]

[2]
[3]
[4]

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Moormann, A., “Analysis of Wall and Ground
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[18]

[19]

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GeotechnicalandGeo-Environmental Eng., ASCE, 2005 pp. 162-177.
Peck, R.B., “Deep Excavation and Tunneling in SoftGround”, In Proc. of the 7th Int. Conf. on Soil
Mechanics and Foundation Eng., Vol. 1, 1969, pp.
225–290.
Clough, G.W., and O’Rourke, T.D., “Construction
Induced Movements of in Situ Walls. In Design and
Performance of Earth Retaining Structures”, Edited by
P.C. Lambe and L.A. Hansen, American Society of
Civil Engineers, Geotechnical Special Publication,
No. 25, 1990, pp. 869-884.
Goldberg, H., “Lateral Support Systems andUnderpinning,
Construction
Methods”,
Rep.
FHWA-RD-75-128, 129 and 130, Federal Highway
Administration, Washington, D.C., 1976.
Hsing, B.-C. B., “A Case Study on the Behaviour of a
Deep Excavation in Sand”, Computers andGeotechnics, Vol. 36, 2009, pp. 665-675.
Brahana, P.E., William, M., Tanner, E.I., and Larry,
D. “Deflection and Earth Pressure Measurements of an
Anchored Concrete Shoring Wall”, 7th Int. Symp. on
Field Measurements in Geomechanics, 2007.
Jalla, R., “Basement Wall Design for GeotechnicalAspects” J. of Architectural Engi., 1999, pp. 89-91.
El Swafe, M., and Nazir, A. K., “The Effect of Deep
Excavation-Induced Lateral Soil Movement on
Behavior of Strip Footing Supported on Reinforced
Sand”, J. of Advanced Research of Cairo University,
Egypt. 2011.
Krasinski, A., and Urban, M., “The results of analysis
of deep excavation walls Using Two Different
Mathods of Calculations”, Archives of Civil Eng.,
Vol. LVII, 2011, No. 1.
Mohamed, A. A., “The Effect of Ground Movement
Induced by Retained Excavation on Adjacent
Structures”, PhD. Thesis Submitted for Examination,
Civil Eng. Dept., Al-Azhar University, Egypt 2011.
PLAXIS Version 8.0 Material Models Manual, “Finite
Element Code for Soil and Rock Analysis”, 1998.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Enhancing Stability of Slopes with Cement Kilin Dust Sand Mixture Layer
Al Serif M. Abd Alazez 1, Ahmed M. Eltohamy 2 and Ahmed Rusdy Towfek 3
1 Professor of Geotechnical Eng., Faculty of Engineering, Assuit Univ., Egypt.
2 Associate Professor of Geotechnical Eng., Faculty of Engineering, Beni Suafe Univ., Egypt.
3 Assistant Professor of Geotechnical Eng., Faculty of Engineering, Azhar Univ., Egypt.

ABSTRACT
Cement kiln dust, CKD is a by-product produced at relatively high rates at cement manufacturing plants. In many
Egyptian cement plants CKD is completely land filled at nearby valleys causing economical and environmental
problems. A numerical prototype dimensio ns model has been experimentally verified and investigated to examine theeffect iveness o f stabilizing sand slopes by a surface layer o f sand-(CKD) mixture. It was concluded that addit ion of(CKD) to sand improves compaction properties, compressive strength and shear strength. A numerical prototype
dimensio ns model has been experimentally verifie d and invest igated to examine the effectiveness of stabilizing sandslopes by a surface layer o f sand-(CKD) mixture. It was concluded that covering the sand slope with a layer ofsand-(CKD) mixture effect ively improves the carrying capacit y o f strip footing loading on slope crust level. The
improvement increases wit h increase of (CKD) addit ion ratio and soil densit y.
Keywords: Cement kiln dust; Enhancing slope stability

1. INTRODUCTION
Cement Kiln Dust (CKD) is one o f the byproducts of cement
industry that is produced in relat ively large quant ities. (CKD)
is a finally divided, dry practice material drained from cementkiln by exhaust gases and captured by the kilns’ air pollution
filt ering system, [1]. The raw input material of cement
industry includes limestone shale, sand and iron. Ratios andtype of these materials together with manufacturing
techno logy, required product specificat ions, kiln type andfuel t ype which determine the chemical co mposit ion andphysical properties of the byproduct [2]–[6]. Appropriate
disposal o f (CKD) is a challenging problem in cement
industry due to environmental concerns. Alt hough modern
advances in cement manufacturing techno logy resulted inconsiderable reduction of the produced quantit ies of (CKD),
however production rates may st ill be considered as anindustry byproduct being produced in relat ively large
quant it ies. The majorit y o f Egypt ian cement manufacturing
plants transport (CKD) to land filling areas with considerable
transportation costs, environmental po llut ion problems andcertainly in an unecono mical way of dealing with thisproblem, [7]. The composit ion of (CKD) is simila r to that of
cement, consist ing of alumina, silica, calcium oxide, alkalis
and sulfates. The high alkalinit y and fine part icle size, in
addit ion to its cement ious properties make this material
usable for several applicat ions, [8]. There are a number ofstudies that indicated the effect iveness of using (CKD) in a
number of geotechnical, civil, structural and agricultural
applicat ions, [9]–[21].

The effeteness of using (CKD) as a stabilizer of granular so il
was studied experimentally by a number of researchers, [22]and [23]. It was concluded that the unconfined compressive
strength of sand-(CKD) mixtures increases as (CKD) in
addit ion ratio increases up to 75-100% by total dry weight. In
other cases peal unconfined string that mixture was achieved
at 15% of (CKD) by total weight, [24]. It was also found thatthe compressive strength increased with the increase of
curing time, [22]. Maximum dry unit weight (MDUW) and
optimum mo isture content (OMC) are affected wit h (CKD)
addit ion to granular soil. This was attributed to the abilit y o f(CKD) to absorb water and alter soil structure through the
pozzoloric and cation exchange reactions, [25]. In a numberof previous studies (MDUW) decreased while (OMC)
increased as (CKD) ratio by dry weight increased, [26]-[28].
Other studies observed a reverse trend, [23] and [24]. The
effect of salts in calcareous soils on long term performance ofsoil stabilizers such as (CKD), may be eliminated by usingasphalt coating and wrapping techniques, [30].
As can be concluded fro m the previously presented review o f
research work, one of the challenging problems that thecement industry faces is to either reduce (CKD) or to explore
addit ional ways for beneficial uses o f the by-product. The
relat ively high lime, alkaline content and high finesses of(CKD) make it a potential candidate for stabilizing soils. In
Egyptian cement manufacturing plants the main production
rate of (CKD) is nearly 2% of the amount of produced
cement. This relat ively high (CKD) production rate motivated
the researcher to examine the effect iveness of using (CKD) instabilizing slopes with a covering layer of (CKD) sandmixture.

154

2. MATERIALS AND EQUIPMENTS
2.1 SandWashed, air dried siliceous yellow sand was used in themodel test. The sand was sieved through sieve No.4 with
opening size of 4.75 mm. The specific gravit y o f soil particles
was determined by the gas jar method. The main value
determined from 3 tests was 2.66. The maximum andminimum dry unit weight of the sand was 18.22 and 15.45
kN/m3, respectively. The grain size distribution wasdetermined using dry sieve method and the result s are shownin Fig 1. From the grain size curve it was concluded that, D 10 ,D 30 and D 60 were 0.3, 0.6 and 1.0 mm. Uniformit ycoefficient, C u and coefficient of curvature, C c were 3.3 and
1.14, respectively.
100
90
80
70
% Passing

60
50
40

distribut ion analysis was carried out on the sample. From thisanalys is it can be concluded that, effective diameter D10=4.0
Mm. D30=7.0Mm, D60=8.0 Mm, Uniformit y coefficient
Cu=2.0 and coefficient of graduation Cc=1.53.
Table (1) Physical properties of sandRelative densit y
80%
Parameter
60%
70%
17.9
Unit weight17
17.4
γ (kN/m3)
Youngs' modulus
8000
10000
12000
E (kN/m2)
0.35
0.3
0.27
Poissons' ratio υ
ο
41
36
37
Friction angle φ ( )Table (2) Oxides composit ion o f tested (CKD)
Mean ratio %
Oxide
SiO 2
10.12
Al 2 O 3
2.0
Fe 2 O 3
4.03
CaO
41.57
MgO
0.57
SO 3
18.08
NaO
2.51
K2 O
7.94
Cl
7.19
LOI
5.26
2.3 Sand-(CKD) mixture

30

2.3.1 Compaction properties of mixture

20
10
0
0.01

0.1

1
Diameter, (mm)10Fig.1 Particle size distribut ion of sand.
The sand can be described as course to medium sand andaccording to unified so il classificat ion system sand can beclassified as (SW). The shear strength parameters of sandwere determined by drained triaxial co mpressio n test withthree different confining pressures under vertical mo vement
rate of 1 m m/min. Table 1 illustrates unit weight, strain andshear strength parameters of sand co mpacted to different
relat ive densit ies o f 60, 70 and 82%.
2.2 Cement kiln dust
Table 2 illustrates ratios of oxides composit io n of CKD
collected along a production day at "CEMAX" Cement
production co mpany, Assuit, Egypt. As can be indicated fro m
the table (CKD) has relat ively high free lime oxide (CaO) of
41.57% and sulphate oxide (SO 3 ) of 18.08%. Thus the
production of cement ious compounds on hydrat ion of (CKD)
was expected. A fresh sample o f (CKD) was put in a plastic
bog and brought to the geotechnical laboratory. Crain sizeStandard compaction tests were carried out to determine
compact ion properties of sand-(CKD) mixture. Fig. 2
illustrates the relat ion between mo isture content and dry
densit y of pure sand and sand-(CKD) mixture for different
addit ion ratios of 10, 20 and 30%. The optimum mo isture
content (OMC) can be observed to increase almo st linearlywith the increase of (CKD) rat io. The recorded values are 7.7,
9, 10.5 and 11.5% for pure sand, (CKD) ratios of 10, 20 and
30%, respectively. The maximum dry densit ies (MDUW)
corresponding to (CKD) addit ion ratios are 19.32, 19.56 and18.8 kN/m3, respectively. Co mparing the maximum dry
densit y of mixtures and that of pure sand an increase inmaximum dry densit y of 10, 11.6 and 7.6% compared to puresand case for 10, 20 and 30 addit ion ratios, respectively. This
general trend of the variation of (OMC) and (MDUW) withincrease of (CKD) ratio agrees with previous research work
as in [26] - [28].
2.3.2 Compressive Strength:
Sand-(CKD) mixture was compacted at maximum dr ydensit y in cubes 100x100x100 mm to measure compressive
strength of cubes wit h varying (CKD) ratios and caring timeof 3, 7, 14, 28 and 56 day. Different mixes Mix (1), Mix(2)and Mix (3) with (CKD) ratio of 10, 20 and 30% were155

investigated. Fig.3 illustrates the average value o fcompressive strength of three specimens prepared of different(CKD) ratios, and curing time. It can be observed that
increasing (CKD) ratio up to 20 and 30% results in increasing
compressive strength of mixture by 16 and 42% as thecorresponding value recorded with (CKD) ratio of 10%. It

10, 20 and 30%. For a larger curing period up to 56 days therecorded compressive strength was 2.7% only of thecorresponding value at 28 days for different (CKD) ratios.
2.3.3 Shear Strength and Stress Strain parameters:
Drained triaxial test was carried out on different adopted
mixtures Mix (1), Mix(2) and Mix (3) to determine stressstrain parameters. Direct shear test was also carried out on
mixtures to determine shear strength parameters. Table (3)illustrates measured parameters of different mixtures.
Table (3) stress – strain and shear parameters of different
mixtures:Mix (1)Mix (2)Mix (3)
Parameter(CKD)10%(CKD)20%(CKD)30%
Unit weight
19.3
19.6
18.8
& (KN/m3)
Young's
31000
40000
52000
modulus E(KN/m2)
Passions
0.26
0.24
0.22
ratio
Friction
26
20
19
angle (°)
Cohesion C
825
40(KN/m2)

Fig. 2 Relation between mo isture content and dry densit y for
different (CKD) ratios of 10, 20 and 30%.

2.4 Sand Container
The sand container mounted under the loading frame isillustrated in Fig. 4.The tank dimensio ns are 500 * 1200mm
with a depth of 1000mm. One side of the tank is made of
Perspex 10mm in thickness, while the other three sides
together with the base made fro m steel sheets 3mm inthickness. The steel sheets were stiffened by steel angles. A
manual lo ading screw jack was used to apply vertical load tothe strip model footing 100mm in width, 50mm depth and
500mm in length ho llow steel sect ion was used. The load wasapplied at two points separated by 250mm. The bottom andsides of the footing was covered by a sand paper to mobilize
sufficient ly the interface between footing and sand. The
manual screw jack was mounted to the bottom flange of theloading frame which was fabricated fro m a steel beam (BFIB
No.20).
3.0 EXPERIMENTAL PROCEDURES AND TESTINGPROGRAM

Fig. 3 Relat ion between curing time and co mpressive strengthof sand-(CKD) mixture of different addit ion ratios of 10, 20
and 30%.
can be observed that the compressive strength of mixture
increases as curing time increases with a relat ively high rate
up to 28 days. The increasing rat ios were 30, 72 and 114% for
curing time of 7, 14 and 28 days, respectively compared to thecorresponding value at 3 days for different (CKD) ratios of

To examine the effect iveness enhancing stabilit y o f sandslopes by a covering layer o f sand-(CKD) mixture thefo llo wing experimental test ing program was adopted. Fig 5
illustrates strip footing slope system in the plain strain
experimental model before and after improvement with acovering sand-(CKD) mixture layer. The parameters adopted
for the testing program are:156

4.0 FINITE ELEMENT ANALYSIS:
4.1 Conditions analyzed:
A series of two dimensio nal finite element analys is (FEA) of
footing slope system model were performed on the sand slopemodel before and after adding the covering stabilizing (CKD)
sand mixture. The (FEA) will serve to assess the laboratorymodel tests and find out deformat ion trends on soil body.
Finite element code for soil and rock analys is plaxis [29]computer soft ware was used in the analys is. The init ial
condit ions include init ial ground water conditions, init ia lgeometry configurat ion and init ial effect ive stress state. As
the sand used in the laboratory model was dry the groundwater table was chosen to be at the bottom level o f the modelas the default of the program for dry so il condit io n. The init ial
stress condition of the slope was first generated by applying
the gravit y force due to soil weight.
4.2 Finite element modeling:

Fig. 4 Sand container mounted under loading frame
i.
ii.
iii.

relat ive densit y o f sand (Dr)=60, 70 and 82%.
addit ion ratios of (CKD) in sand-(CKD) mixtureof 10, 20 and 30%.
distance of strip footing fro m slope b/B = 0, 1 and
2.

The software program enables automat ic generation of 15
mode triangle plain strain elements for the soil and thefooting. Displacements are presented in the form of a fourthorder interpolat ion funct ion. The element stiffness matrix isevaluated by numerical t itration using twelve stress po ints.
The nonlinear Mohr Co lumn criteria were adopted to model
the soil. The geo metry o f typical finite element model
verified for the analys is is shown in Fig.6. The parameters
used in the (FEA) were determined fro m experimental tests
carried on sand and) sand-(CKD) mixture as previously
presented. Different finite element mesh coarseness were
investigated to decide which to give the results most close to
the experimental model.

a- Before improvement

Fig. 6 Geometry of finite element model verified for the
analys is.
b- After improvement
Fig. 5 Strip footing slope system before and after
improvement with sand-(CKD) mixture layer.

A refined mesh was adopted to eliminate the mesh coarseness
dependency of the finite element modeling for cases of
unstabilized and stabilized slopes. Since the ground surface ofthe slope is not horizontal. A predisceribed footing load wasthen applied in increments with iterat ive analysis up to
failure. To model the boundary condit io ns of the157experimental testing program, the vertical boundary is freevert ically and constrained horizontally while the bottom
boundary is fully fixed. After co mparing the results of theexperimental model and the (FEA) study on model scale, thecomputer program so ft ware PLAXIS was used to investigate
the affection o f selected parameters on prototype scale
numerical model.

indicated. Similar trends have been indicated for other casesof (b/B) = 0 and 2, however curves corresponding to suchcases will not be presented for abbreviation. Fig. 8 indicatesbearing capacit y settlement curves for the stabilized slope
case wit h sand-(CKD) mixture of 10% addit io n rat io of
medium sand.
qu (kN/m2)
0

5.0 RESULTS AND DISCUSSION:

100

150

150

200

250

300

1

200

Settlement (mm)Bearing capacit y settlement curves were sketched for
different (b/B) ratios and different soil relat ive densit ies forexperimental and numerical models. Fig. 7 illustrates bearing
capacit y settlement curves for (b/B=1.0) for medium so il
densit y so il as an example to examine the adopted numericalmodeling.
qu (kN/m2)
50

100

0

5.1 Verification of numerical model:

0

50

250

2
3
4
5
6

0

Num.
Exp.

7

Settlement (mm)1Fig. 7 Bearing capacit y settlement relat ion for medium sand(b/B= 1) for stabilized slope case (CKD) = 10%.

2
3
4
5
6

Num.
Exp.

Fig. 7 Bearing capacit y settlement relat ion for medium sand(b/B= 1).
As can be observed fro m this figure the bearing capacit ysettlement carvers for excremental analys is can be divided tothree stages. The first stage is approximately a straight line,
where the soil behavior is mainly elast ic. This stage ends atbearing capacit y equaling 0.4 of ult imate value (qu) andsettlement equaling 0.2 of ult imate settlement (Su). The
second stage is of a carved trend where so il starts to plast ify.
In this stage bearing capacit y increases wit h the increase of
settlement up to the ultimate load at which the tangent to thecurve beco mes horizontal, this corresponds to settlement
equaling 5% of footing breadth. In the third stage loaddecreases rapidly in a non linear trend with the increase of
settlement with a plast ic so il performance. The marked peakvalue indicated with experimental curves indicatesoccurrence of general shear failure for different soil densit ies
due to existence of slope. The previously described trend wasin good agreement with the previous studies, [30], [31] and[32]. Curves corresponding to numerical investigationreflected a close trend to experimental curves at the init ial
stage with slight increase of modulus of elasticity of so il.
Inspite of the relative close ult imate bearing capacit y ofexperimental and numerical carves no marked peak value wasmedium and loose sand cases, respect ively. As can beindicated from these figures curves corresponding for thestabilized slope case preserved the same trend for bothexperimental and numerical cases as that of unstabilized
slope. The same trend has been also observed with other cases
including (b/B= 0.2) and different (CKD) addition ratios of
10 and 30%. The ult imate bearing capacit y o f experimental
analys is overest imated that of numerical analys is by average
percentage ratios of 0.05 and 0.15% for unstabilized andstabilized slopes, respect ively. Hence, it can be concluded
that numerical analys is, with the adopted mesh coarsenessand condit ions results in a fairly good agreement of stressstrain behavior of footing rest ing stabilized and unstabilized
slope. This applies also for the relatively close ultimate
bearing capacit y values. The deformed so il profile underneaththe strip footing at failure is illustrated in Fig. 9.
5.2 Prototype scale numerical model:
Figs. 10, 11 and 12 illustrate the relat ion between the bearing
capacit y of the strip footing and the addit io n ratio of the(CKD) stabilizer with different densit ies of loose, mediu mand dense sand for relat ive footing distance (b/B) of 0, 1 and
2 , respectively. For the case of the strip footing located at thecrust of slope, stabilizing the slope surface layer wit h (CKD)
of 10, 20 and 30% addit ion ratio results in enhancing bearing
capacit y by 22, 46 and 67%, respectively for dense sand case.
For medium sand the recorded ratios were 24, 34 and 48,
respectively.

158

Fig. 9 Deformed shape profile of soil under the footing at
failure.

Fig. 9 Effect of (CKD) addition ratio on bearing capacit y offooting for different soil densit y for (b/B =0).

Fig. 11 Effect of (CKD) addit ion ratio on bearing capacit y o ffooting for different soil densit y for (b/B = 2).
Stabilizing the surface layer for the case o f loose sandappeared to have a smaller effect as the enhancement ratios
were 20, 30 and 40%, respectively. Slightly higher
enhancement ratios have been observed with the strip footing
shifted away fro m the slope crust, (b/B=1). Enhancement
ratios for dense sand were 37, 47 and 69% for 10, 20 and 30%(CKD) addit ion ratios, respectively. Further shift ing of the
strip footing fro m slope crust (b/B=2) does not result inincreasing bearing capacit y co mpared to the more close caseof (b/B=1.0). Hence it may be stated that the suggested
technique appears to be more effective in enhancing thecarrying capacit y o f strip footing for the case of footing
distance from crust of more than footing breadth. Slight ly
lower enhancement in carrying capacit y has been recorded as
the footing is more close to the slope crust. Figs. 12 and 13
indicate total displacement at failure of a slope loaded withstrip footing before and after stabilizat ion wit h sand-(CKD)
mixture, respectively. As can be indicated from these figures
the enhancement of carrying capacit y o f a strip footing ma y
be attributed to that the stabilized layer with relat ively higher
unit weight and enhanced engineering properties results indirect ing the lateral component of the foundation so il
settlement under the strip footing to the downward direction,
the matter that results in not affect ing the stabilit y of the slope
and hence increasing the caring capacit y o f the footing
compared to the unstabilized slope. For the relat ively clo se
strip footing to the slope for the foundat ion so il to deform andhence the footing to settle down, the deforming so il has to lift
up and push beside the stabilized so il layer. The relat ively
high unit weight of stabilized layer will hence resist the
deformation o f soil and settlement of foundat ion.

Fig. 10 Effect of (CKD) addit ion ratio on bearing capacit y o ffooting for different soil densit y for (b/B = 1).

159

(CKD) addit ion ratio and soil densit y. The suggested
technique appears to be more effective in enhancing thecarrying capacit y strip footing for the case of footing distance
fro m crust was more than footing breadth. Slight ly lower
enhancement in carrying capacit y has been recorded as thefooting is more close to the slope crust.
7. REFRENCES

Fig. (12) Deformed numerical so il profile of unstabilized
slope for (b/B =1).

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Fig. (13) Deformed numerical so il profile of (CKD) surface
layer stabilized slope for (b/B =1).
6.0 CONCLUSIONS
Cement Kiln Dust (CKD) is one o f the byproducts of cement
industry that is produced in relatively large quant it ies. In
Egyptian cement manufacturing plants the main production
rate of (CKD) is nearly 2% of the amount of produced
cement. The produced (CKD) has high lime, alkaline content
and high finesses which make it a potential candidate forstabilizing so ils. It was concluded that addit ion of (CKD) tosand improves co mpact ion properties, compressive strengthand shear strength. A numerical prototype dimensions model
has been experimentally verified and invest igated to examine
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[21] Santagata, M. C., and Bobet, A. "The use of Cement Kiln
Dust (CKD) for Subgrade Stabilization/Modification"(No. SPR 2575), Purdue University,West Lafayette,
Indiana, USA, 2002.
[22] Baghdadi, Z. A., Fatani, M. N., and Sabban, N. A., "Soil
Modification by Cement Kiln Dust", Journal of
Materials in Civil Engineering, Nov. 1995, pp. 218-222.
[23] Baghdadi, Z. A., and Rahman, M. A, "Potential of
Cement kiln dust for the stabilization of dune sand in
highway construction", Building and Environment,
25(4), 1990, pp. 285-289.
[24] Taha, R., "Evaluation of Cement Kiln Dust-Stabilized
Reclaimed Asphalt Pavement Aggregate Systems in
Road Bases", Transportation Research Record, Journalof the Transportation Research Board, TRB Record
1819, Washington D.C., USA, 2003, pp. 11-17.
[25] Bhatty, J. I., "Alternative Uses of Cement Kiln Dust,
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PCA R&D 327, Portland Cement" Association,
Skokie, Illinois, 2003.
[26] Zaman,M., Laguros,J.G., Sayah.A. "Soil Stabilization
Using Cement Kiln Dust", Proceedings 7thInternational Conference on Expansive Soils. Dallas,
Texas, 1992, pp. 347-351.
[27] Miller, G.A. and Azad.S, "Influence of Soil Type on
Stabilization With Cement Kiln Dust", Construction
and Building Materials, 14(2), pp. 2000, 89-97.
[28] Sreekrishnavilasam, A.; Rahardja, R.; Kmetz, R., andSantagata, M., "Soil Treatment Using Fresh andLandfilled Cement Kiln Dust", Construction and
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[29] Bringkgreve, R.B.J, Vemeer, P.A., PLAXIS – finite
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[30] Zadroga, B 'Bearing capacity of shallow foundation
on noncohasive soils", Journal of Geotechnical Eng.,
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capacity of strip footing on geogrid-reinforced sandslope" Geotextile and Geomembrane, 19, 2001, pp.
279-298.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The effect of confining stress on the analysis of excavations adjacent to existingbuildings
Sabzi Zahra1, Fakher Ali2.
1: PhD student; School of Civil Engineering; Faculty of Engineering; University of Tehran; Iran
2: Associate Professor, School of Civil Engineering; Faculty of Engineering; University of Tehran; Iran
ABSTRACTIn the design of urban excavations controlling soil and adjacent buildings’ displacement are important. The accurate
prediction of displacements is difficult because they are influenced by several parameters. Soil stiffness is an
important parameter in design and it is often assumed as a constant value, but stress distribution in the soil changes
during the excavation and consequently changes the stiffness.
A vast zone of Tehran is composed of coarse-grained cemented sediments and its stiffness is highly influenced byconfining pressure. In this article, a verified FE analysis has been performed to investigate Tehran soil behavior in
excavation. The used soil model is a nonlinear elastic- plastic model in which the effect of confining pressure on
stiffness has been indicated using Janbu model that was calibrated for Tehran soil.
Results show that assuming stiffness as previously mentioned, influences the estimation of wall displacement, groundsettlement and also the type of ground support.
Keywords: excavation, displacement, Tehran soil, stiffness, confining stress.

1. INTRODUCTION
In urban areas, the design of new buildings often
necessitates the construction of braced excavations. During
excavation, the state of stress in soil changes . These changes
in stress are typically expressed in the form of lateral
movement and settlement of the ground which can havedamaging effects on adjacent buildings.
Shoring and underpinning are methods used as supports in
excavations. Shoring is a form of temporary support which
can be tie to existing buildings adjacent to excavation to avoid
damages. Underpinning is a temporary support that its mainfunction is to transfer the load carried by a foundation from its
existing bearing level to a new level at a lower depth. [1]Similarly, one type of shoring method, called “Inclined
Strut”, is widely used in different areas of Iran and isconsidered a traditional method in Tehran (Fig1). Inclined
Strut can be considered as shoring or underpinning.
In the design of braced excavation the target is to minimize
the cost of the support system while ensuring that the support
is not so flexible that it causes additional soil displacementthat damage adjacent buildings. The design of braced
excavations is a complicated soil-structure interaction
problem. Usage of the numerical modeling to analyze thesesoil-structure interaction problems by engineers is increasing.
Many investigators [2], [3], [4], [5] have verified thereliability of FEM method in analyses of excavations.
In addition to reasonably modeling the actual construction
sequences, selecting an appropriate soil model and thecorresponding soil parameters is an essential step in the FEM
analysis of excavation problems. Soil stiffness is an importantparameter that is often assumed as a constant value but
actually during excavation, soil stiffness changes. Previous
studies [6] have demonstrated that the soil stiffness in Tehran
at small-strain can be higher than stiffness of most gravelly
soils reported in technical literature.
In this paper, a verified FEM analysis has been conducted toinvestigate Tehran soil behavior, considering the effects
stiffness changes based on Janbu model.[7]Fig 1: schematic traditional shoring method
2. EFFECT OF CONFINING STRESS ON SOIL
BEHAVIOR
Elastic stress-strain relations are introduced with twoconstants which can be bulk modulus K and shear modulus G.
In nonlinear elastic models, when stress state changes,
elastic parameters K, G change too. Comprehensive
description of a model requires presenting changes of K, G in
relation to changes in stress states. [6]Constitutive models indicate stress-strain relationships.
The key to a successful design is to use appropriate

162

parameters of soil models, which are able to present real
behavior of soils. Description of nonlinear stress-strain
behavior is necessary in excavation analysis. Many models
have been presented, in which, soils have taken nonlinear
elastic materials. Stiffness is one of the important soil
parameters and is a variable that relates axial or shear stressincrements to axial or shear strain increments. Stiffness varies
with strain increase in which it has the most value when strain
is small and decreases when strain increases. So assuming
stiffness as a constant value is wrong. Nonlinear models of
soils include two separate questions. One is determining
stiffness in small strains and the other is ascertaining
relationship between maximum stiffness and stiffness inlarger strains [7]. Nonlinear stress strain curve can be
determined by several methods. Some of these methods
proposed by Duncan J.M. and Chang C.Y. [9], Fahey andCarter [8], Lehane B. and Fahey M. [10], Ramberg-Ozgood,
Seed and Idriss [11] and Janbu N [7]. In this article janbu
model used to simulate behavior soil of Tehran.
Janbu model:
In nonlinear models stiffness can considered dependent on
confining pressure. The effect of confining pressure on
stiffness can be shown using Janbu model (Janbu N, 1963):

σ 'm nG = K G Pa (
)
Pa

(1)

In this equation (Eq. (1)), G is stiffness, σ΄m is average ofeffective stress, Pa is reference stress to normalize theequation (usually Pa= 100 kPa) and n and K G are equation
parameters.

several points of Tehran. By review of other coarse-grainedsoils stiffness, it was observed that the behavior of gravelly
soils with high percentage of gravel, over consolidated
coarse-grained soils, pre-strained coarse-grained soils andcemented soils are similar to Tehran soil. The specificbehavior of Tehran soil and its dissimilarity to othercoarse-grained soils is relevant to its cementation and overconsolidation [6].
The investigations on Tehran soil has been done by
Sangtarashha et al, then Janbu model has been applied
supposing soil behavior [6]. Taking undisturbed samples fortriaxial tests or using self-boring pressuremeter was notpossible at coarse-grained soils of Tehran. Therefore,
Pressuremeter test has been performed with stress control
according to method A of ASTM D4719-1994 and parameterof Janbu model obtain for some points of Tehran and some ofthat results used here are shown in table I.
Table I: Janbu models parameter obtained for Tehran sediments

Deposit Type
A
B
C

KG
16500
7120
4870

n
0.6
0.5
0.6

γ (kg/m3)
2200
1850
2000

Studies about Tehran soil indicate that stiffness increase
with depth has similar trend in different points of Tehran andjanbu model can estimate it well. Coefficient n in this Eq.(1)varies between 0.5 and 0.6 for Tehran soil and in most points
is about 0.6. [6]
4. NUMERICAL MODELING

3. GEOLOGY AND CHARACTERIZATION OF
TEHRAN SOIL
Alluvial deposits of Tehran have been studied by many
researchers during recent years. Numerous classifications ofthese deposits have been proposed. Observations of Tehran
alluvia have shown that grain size and shape, sedimentary age,cementation, grain contact type, faults, fractures andweathering affect the geotechnical properties of thesecoarse-grained materials [12]. The effects of these factors
have been discussed by Fakher [13]. Rieben [14] divided the
Tehran alluvia into four categories based on geologicalfactors, and identified them as A, B, C and D alluvia. The A
alluvium is the oldest and D alluvium is the youngest. The
grain sizes of the alluvia are variable and some are naturally
cemented. The cementation grades vary from strongly
cemented to noncemented depending mainly on the nature of
the cementitious material. Strongly cemented Tehran
alluvium is usually seen in A alluvium and is a secondary
development resulting from the recystallization of previously
dissolved calcite carbonate of the source rock. Moderately
cemented Tehran alluvium is usually seen in C alluvium andis clayey. The B alluvium presents weak and variable
cementation, and D alluvium, being the youngest, has no
cementation. A unique trend was observed in increase of soil
stiffness with in relation to depth or confining pressure at

In this research, we performed two-dimensional total stresselasto-plastic FEM analysis to examine ground movement
and stress in the ground after installation of inclined struts in
excavation.
Two-dimensional numerical simulations were carried out
using a FEM analysis code. Fig. 2 shows the FEM mesh. A
large zone was selected to avoid any measurable effects fromthe boundary in the final results. In this research, we use
rectangular elements for two-dimensional analysis. Theseelements have four nodes and four integration points, and theGaussian integration method was applied to them. The
inclined struts and neighboring building were modeled using
beam elements. The analyzed soil has 861 nodes and 800
elements in total. Mohr-Coulomb constitutive model waschosen for soil elements. Dimensions and soil parameters ofhypothetical excavation case are shown in Table II. To
minimize boundary effects, the vertical boundary at the far
ends was set almost as 3 times of excavation’s width from thecenter of excavation. Therefore, it was assumed to be free invertical direction and restricted in horizontal direction. The
bottom horizontal boundary was restricted in both horizontal
and vertical directions. The boundary condition of model isshown in Fig.2 too.

163

A series of numerical analysis was performed toinvestigate the effect of confining stress on selecting thestiffness of soil and consequently on the deformation causedby excavation. Fakher et al have been done the studies todetermine geotechnical properties based on geologicalfeatures. Their proposed stiffness parameter for A, B and C
alluvia are shows in Table 3.[13] Linear analyses have been
done on the A, B and C alluvia with parameters presented inTable III: Geotechnical parameter obtained for Tehran sediments

Deposit TypeA
B
C

Fig.2. Finite element mesh used for the hypothetical excavationcase.

In modeling stages, firstly, the in-situ horizontal andvertical stresses were generated. Initial in-situ horizontal andvertical stresses are as follows.
σ z = γhσ x = K 0σ z

Which γ is the soil density, K 0 is the coefficient of earthpressure at-rest, and σ z , σ x are the vertical and horizontal
initial stresses at depth of h respectively. Then, it was
assumed that the building was located next to the excavation
and applied to model and excavation stages are modeledaccording to common excavation procedures. Finally inclined
strut added to model as a support system.

ParameterC(kN/m2)φ(degree)

H
(m)

B
(m)

E steelE concreteamountTable II: The parameters of soil and struts used in numericalmodeling

75

30

5

10

2.0E+82.0E+7E (MPa)200
100
50

Variation of stiffness with confining stress based on Janbu
relation are shown in Fig.3 for A, B and C alluvia using theparameters presented in Table 1. To investigate the effect ofconfining stress on the stiffness, Janbu model applied toconstitutive relations and soil stiffness considered as asolution dependent parameter. Calculating the confining
stress in each step, related stiffness calculated and applied toanalysis.
Results of analyses are illustrated in Fig.4-Fig.6
Fig 4, 5 and 6. show the comparison between the results ofanalysis with considering stiffness as a constant value and theresults of analysis with considering variation of stiffness withconfining stress. Results show that the wall deflection andsurface settlement obtained from nonlinear analysis (stiffness
is solution dependent) are significantly more than obtained
from linear analysis (stiffness is a constant value).

Fig 4. a) Deformation of excavation wall. b) Ground surface settlement. c) Ground surface lateral displacement for soil A
164

Fig 5. a) Deformation of excavation wall. b) Ground surface settlement. c) Ground surface lateral displacement for soil B

Fig 6. a) Deformation of excavation wall. b) Ground surface settlement. c) Ground surface lateral displacement for soil C
Excavation leads to reduction in lateral pressure in someelements and consequently confining stress decrease andrelative stiffness decrease too and if only a constant modulus
of elasticity is considered i.e., the degradation of stiffness due
to excavation is ignored, the displacements are obviously
underestimated.
Results show that loads in inclined strut obtained fromlinear analysis is less than obtained from nonlinear analysis.
Considering stiffness as a constant parameter cause tounderestimate the strut loads and then lead to underestimate
the design loads and consequently cross section of support
system.

5. CONCLUSIONS

165

1. Stiffness of soil of Tehran increases with depth andconfining stress and significantly decreases withexcavation.
2. In order to reasonably modeling the excavation and tounderstanding the actual soil behavior, selecting anappropriate soil parameters is an essential step in theanalysis. Considering variation of stiffness withvariation of confining stress can rationally model the
excavations.

3. The results of the analysis with constant stiffness ofsoil show that the excavation-induced deformation
can be underestimated and serious damage can occur
in real behavior of soil.
4. These result show that considering soil with greater
stiffness than real stiffness cause to underestimate theloads in inclined strut.

zone. Quarterly Journal of Engineering Geology andHydrogeology, 2007. 40, 163–174.
[14] Reiban, E.H. “Geological Observation on Alluvial
Deposits in Northern Iran”. Geological Organization of
Iran, 1966. Report, 2 [in French].

6. REFERENCES[1] Sadeghian S, Fakher A, “An Investigation into a Shoring
Method to Support Buildings Adjacent to Excavations”
The 17th Southeast Asian Geotechnical Conference
proceeding., (2010)[2] Boone, S. J., “Design of Deep Excavation in Urban
Environments”, Ph.D. thesis, Univ. of Toronto. (2003).
[3] Kung, G. T. C., Juang, C. H, Ou, C. Y., (2009),“Modeling Small-Strain Behavior of Taipei Clays for
Finite Element Analysis of Braced Excavations”,
Computers and Geotechnics, Vol 36 (2009), pp 304–319.
[4] Hsiao, C. L., “Wall and Ground Movements in a Braced
Excavation in Clays and Serviceability Reliability ofAdjacent Buildings”, (2007). Ph.D. thesis, Clemson
University.
[5] Tang, Y.G. and Kung, G. T.C. “Investigating the Effectof Soil Models on Deformations Caused by Braced
Excavations through an Inverse-Analysis Technique”,
Computers and Geotechnics, Vol 37 (2010) pp769–780.
[6] Sangtarashha K. Fakher A. Pahlevan B. “Variation of
stiffness of Tehran coarse-grained soil with depth andstrain”[7] Janbu N, “soil compressibility as determined byoedometer and triaxial tests.” Proc. Europen Con. on Soil
Mech. Fndn. Engng., Weisbaden1(1963)., pp. 19-24[8] Fahey M. and Carter J.P., “a finite element study of thepressuremeter test in sand using a nonlinear elastic
plastic model.” Canadian Geotechnical Journal, (1993).
vol. 30, pp. 348-362[9] Duncan J.M. and Chang C.Y., “nonlinear analysis ofstress and strain in soils.” Journal of the Soil Mechanics
and Foundations Division, ASCE, vol.96 SM5, (1970).
pp. 1629-1653[10] Lehane B. and Fahey M. “a simplified nonlinear
settlement prediction model for foundations on sand.”Canadian Geotechnical Journal, , (2002). vol.39, pp.
293-303.
[11] Seed H.B. and Idriss I.M. “soil moduli and damping
factors for dynamic response analysis.” Earthquake
engineering research center, University of California,
Berkeley, (1970). report EERC 70-10.
[12] A. Cheshomi, A. Fakher and C.J.F.P. Jones. “Acorrelation between friction angle and particle shape
metrics in Quaternary coarse alluvia Quarterly” Journalof Engineering Geology and Hydrogeology 2009; v. 42;
p. 145-155[13] Fakher, A., Cheshomi, A. & Khamechian, M. “Theaddition of geotechnical properties to a geologicalclassification” of coarse grain alluvium in a pediment

166

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Bored Pile Socket in Erratic Phyllite of Tuang Formation
D.E.L. Ong1
Swinburne University of Technology (Sarawak Campus), Malaysia

1

ABSTRACTThis paper highlights a case study where about 950 nos. of cast in-situ bored piles were used as a deep foundation
system to support a large manufacturing plant in suburban Kuching, Sarawak, Malaysia. Localised moderately-strong
metamorphic metagreywacke and extrusive igneous andesite are found occurring together with the more dominant but
erratic, metamorphic phyllite due to regional metamorphism. Such combination of geology requires stringent quality
control during construction to ensure sufficient bored pile socketing lengths into the mostly erratic phyllite. For thisvery reason, 33 nos. of Pile Dynamic Analyzer (PDA) and 4 nos. of Maintained Load tests were carried out to determine
the ultimate static capacities of the installed bored piles. The PDA and Maintained Load test results show that all thetested piles meet the pre-determined design working loads.
Keywords: phyllite, socket, Tuang formation, point load test

1 INTRODUCTION
The project location has been found to be influenced bycontrasting geology that consists of the Padawan (mainly
sedimentary), Tuang (mainly metamorphic) and Quaternaryformations. Localised moderately-strong metamorphic
metagreywacke and extrusive igneous andesite are foundoccurring together with the more dominant but erratic
metamorphic phyllite due to regional metamorphism. The
Tuang formation has been noted to be most prominent during
construction, evidenced by the abundant encounters with thefriable and erratic phyllite. The construction of the large
manufacturing plant required thorough understanding of theunderlying geology. A cast in-situ deep foundation system
was recommended for the plant due to heavy imposed loads
and limited tolerance to differential settlement. For thefoundation system to be effective, sufficient socketing length
is required into the erratic phyllite. This paper is an extensionof Ong & Choo (2011) [1].

Recovered cores of phyllite mostly revealed RQD values of
0%, up to a maximum of 75%. Fig. 1 shows phyllite coreswith Total Core Recovery (TCR) of 100% but having RockQuality Designation (RQD) of 0%. This observation isconsistent with indications reported in Tan (1993) [2]. If
RQD is used as a measure of rock strength for design purpose,then literally phyllite has zero strength! However, as ahomogenous rock mass underground, it is impossible thatphyllite has no strength!

2 SITE GEOLOGY
The proposed deep foundation system was socketed ingeology comprising mostly of moderately weathered phyllite,
and flanked at the site boundaries with intrusions ofmoderately to completely weathered shale and strong
moderately weathered andesite. These geological conditions
are synonymous with the Tuang Formation, and to a lesser
extent the Padawan and Quaternary formations. The
manifestation of the metamorphic phyllite rock in the Tuang
Formation is characterised by semi-schistose features, further
evidenced by the parallel arrangements of distinct cleavages
and grains, occasionally interbedded with discrete bands ofquartz and calcite.

Fig. 1. Discontinuous and highly fractured cores of phyllite
3 BEHAVIOUR OF ERRATIC PHYLLITE
An overview of RQD and the Unconfined Compressive
Strength (UCS) tests on the phyllite rock cores is presented inFig. 2. Due to the relatively low RQD values, it was only
possible to conduct UCS tests on 14 of the 194 core runs
extracted (12.4%). A conservative approach for the design ofdeep foundations in rock considers that the pile capacity is
attributed only to the rock strength. Hence, it is critical todetermine the socketing length required for piled foundation
systems. The requirements of the manufacturing plant also167

warranted the use of a piled foundation system to mitigate
serviceability issues, such as differential settlement and pile
movements. Given such conditions, the erratic phyllite
behaviour provides serious challenges to designers.

irregular lumps. Requirements for sample size restricted
testing to be performed on lumps with aspect ratios illustrated
in Fig. 3. Irregularly-shaped test samples for verification
purposes were collected from spoils during boring. Fig. 4
shows rock samples collected from various coring depths,
which were subsequently collected for point load testings on
site.

Fig. 4. Excavation spoils collected from various boring
depths.

Fig. 2. Overview of RQD and UCS tests on phyllite cores(per 1m run)
3.1 Point load test on site
The outcome of the soil investigation works signalled
expected difficulties in verification of socketing material
during the construction of the deep foundation system. The
occurrence of phyllite as well as its RQD values has been
observed to be very low, most likely due to its friableness. As
such, point load testing was suggested to be used as acomplementary index test to help correlate the UCS values
tested on the limited and unpredictable solid cores of phyllite.

Fig. 3. Shape proportion for point load testing on irregularlumps (AS 4133.4.1-2007).
The point load testing apparatus was deemed suitable for fielduse due to its simple operation. The procedures outlined in AS
4133.4.1-2007 [3] were preferred for carrying out point load
testing due to the allowances for tests to be carried out on

3.2 Review of test results
Table 1 shows the summary of results for 3 types of tests
carried out as an attempt to characterise phyllite, namely (i)Point Load Index test on lump samples, (ii) Point Load Index
test on rock core (axial direction) and (iii) UnconfinedCompression Strength (UCS) test. It can be seen that the
average Point Load Index, I s50 value for lump samples within
depths of 10.0m-11.9m, 12.0m-13.9m, 14.0m-15.9m and16.0-17.9m are 3.094MPa, 3.749MPa, 3.866MPa and3.654MPa, respectively. No lump samples were available fordepths exceeding 18m as the preliminary design did not
warrant any bored pile socket length greater than 18m.
Despite the consistency of the average values over the various
depths as reported, the standard deviations are rather poor i.e.
at 2.033MPa, 2.838MPa, 2.696MPa and 2.510MPa,
respectively. In other words, if assumed on a normal
distribution probability density function, the probability ofgetting the average I s50 value of 3.753MPa over the 374
samples is only about 15.1%.
Table 1 further shows the difficulty of obtaining intact coresamples (minimum 14.1mm in length) for Point Load test
(axial direction) between depths 10m and 14m due to thefriableness of the phyllite at the upper depths. Between depthsof 14.0m-15.9m, 16.0-17.9m and 20.0m-21.9m, the average
I s50 values for intact cores tested on axial direction are
0.745MPa, 1.235MPa and 1.256MPa, respectively. The
standard deviations recorded are 0.120MPa, 0.12MPa and0.730MPa, respectively. If interpreted on a normal
distribution probability density function, the probability ofgetting the average I s50 value of 1.138MPa over the limited
number of 9 samples is about 43%.

168

Since UCS test requires intact rock core of at least 94mm inlength, relatively competent phyllite is only available from
16m depth as shown in Table 1. At depths of 16m-17.9m,
18m-19.9m 20-21.9m, 22m-23.9m and 26m-27.9m theaverage UCS values obtained are 9.53MPa, 4.45MPa,
5.2MPa, 9.33MPa and 12.06MPa, respectively. The standard
deviations recorded are 2.051MPa, 3.286MPa, 4.030MPa,
5.629MPa and 17.519MPa, respectively. If interpreted on anormal distribution probability density function, theprobability of getting the average UCS value of 8.259MPa
over the 14 samples is only about 4%.
Therefore, surprisingly as a summary on testing in this study,
it seems that the most reliable way of statistically measuring
the phyllite rock strength is via the Point Load (axial core) as
compared to Point Load lump and UCS testing methods.

suggested by Williams et al. (1980) [4] and recommended foruse by Ting and Fuad (2003) [5] for metamorphosed
sedimentary rocks comprising a sequence of slate and phylliteintercalated rhythmically with greywacke sandstone found at
the Batang Rajang Bridge site in Sarawak. In the absence of amore systematic design procedure for phyllite, Ting & Fuad
(2003) [5] suggestion is hence adopted for use in this study as
it is perhaps the closest empirical representation to phyllite
for this case study.
During the design stage when only UCS and I s50 (axial core)values were available, based on judgment and literature
review of Basu & Kamran (2010) [6] and Ting & Fuad (2003)[5], a preliminary design USC value of 7.4MPa is assumed
and this value must be verified through pile load tests
performed on sacrificial piles on site. A factor of safety of 2.0
has been used in the design of the bored piles.

Table 1. Summary of test results based on UCS and Point Load test on phyllite
Information available only during design stage
PLT (axial core)
Depthswhere
phyllite aretestedNo. ofsamplesDepth
10m-11.9m
Depth
12m-13.9m
Depth
14m-15.9m
Depth
16m-17.9m
Depth
18m-19.9m
Depth
20m-21.9m
Depth
22m-23.9m
Depth
24m-25.9m
Depth
26m-28.0m
Over entire
depths10m-28m

Information available only during
construction stage

UCS

Ave.
I s50
(axialcore)(MPa)

Std dev(axial core)

0

N/A

N/A

0

N/A

2

No. ofsamples

PLT (lump sample)

Ave.
UCS

Std devon UCS

No. ofsamplesLump
for I s50

Ave.
I s50
(lump
sample)(MPa)

Std dev(lumpsample)

(MPa)

(MPa)

0

N/A

N/A

23

3.094

2.033

N/A

0

N/A

N/A

77

3.749

2.838

0.745

0.120

0

N/A

N/A

166

3.866

2.696

2

1.235

0.120

2

9.53

2.051

81

3.654

2.510

0

N/A

N/A

2

4.45

3.286

0

N/A

N/A

5

1.256

0.730

3

5.20

4.030

0

N/A

N/A

0

N/A

N/A

0

N/A

N/A

0

N/A

N/A

0

N/A

N/A

2

9.33

5.629

0

N/A

N/A

0

N/A

N/A

5

12.06

17.519

0

N/A

N/A

9

1.138

0.476

14

8.259

9.975

347

3.753

2.637

(MPa)

(MPa)

4 BORED PILE DESIGN

5 PILE LOAD TEST

It is common practice to determine the unit shaft and unit baseresistances based on UCS or q u values of intact rock coresamples, for example, Williams et al. (1980) [3] states that:f s = 0.05q u (MN/m2)(1)
f b = 0.5q u (MN/m2)(2)

In order to improve the confidence level of the designer inview of the difficulty of determining a feasible rock strength
value for phyllite during the preliminary design stage, pileload tests during the construction stage become very
important. The field testing result would then provide data forback-analysis purpose and to subsequently firm up the design
value for phyllite. As a procedure to ensure that the bored pile
can perform satisfactorily at its allowable load limit and toverify the preliminary design UCS value of 7.4MPa, 33 sets

Typical limiting resistance values (ultimate), but may vary
subject to load test results. The analyses carried out in thesubsequent sections are based on the empirical equations

169

of Pile Dynamic Analyzer (PDA) tests as well as 4 sets ofConstant Rate Penetration (CRP) and Maintained Load Test
(MLT) tests were performed on site.
Table 2. shows the results of the 33 Nos. of PDA tests
performed on site. Based on the field test results, the averagefactors of safety provided for the 600mm diameter bored pile
Types 1 and 2, as well as 750mm, 1000mm and 1200mm
diameter bored piles are 2.23, 2.17, 2.06, 2.37 and 2.26,
respectively. These values are in general comparable to thefactor of safety of 2 used in the preliminary design stage.

From the Maintained Load Test results as shown in Table 3,
the magnitudes of the settlements that govern theserviceability limit state design are all less than the limit of
12.5mm, 38.0mm and 6.5mm at Working Load, twice
Working Load and residual conditions, respectively. Thismeans the design UCS ultimate value of 7.4MPa for phyllite
is thus broadly verified in terms of strength and serviceability.
Once sufficient confidence had been achieved on site after
performing the CRP and MLT tests, mass installation ofbored piles commenced.
Table 3. Summary of measured pile head settlement

Table 2. Summary of pile PDA test results
Bored pile
diameter

(mm)

Rock
socketlengthinphyllite(m)

600 (Type 1)
WL=1540kN

2.0

600 (Type 2)
WL=2500kN

2.1

750
WL=3250kN

2.2

1000
WL=4500kN

2.3

1200
WL=5800kN

2.5

Test
pile ID

Measured
PDA ult.
staticresistanceDesignworking
load
(WL)

FOS

(-)

(kN)

(kN)

(-)

23
39
91
106
496
505
511
515
621
779
620
622
624
626
629
633
704
705
735
798
TP1
5
11
13
780
56
508
508
520
778
56
436
470

3,258
3,541
3,310
4,197
3,411
3,277
3,359
3,502
3,257
3,380
5,398
5,607
5,183
5,738
5,349
5,475
5,390
5,283
5,135
5,223
5,875
6,676
6,891
6,588
6,657
9,257
10,553
10,553
11,976
10,9889,257
13,605
12,554

1,540
1,540
1,540
1,540
1,540
1,540
1,540
1,540
1,540
1,540
2,500
2,500
2,500
2,500
2,500
2,500
2,500
2,500
2,500
2,500
2,500
3,250
3,250
3,250
3,250
4,500
4,500
4,500
4,500
4,500
4,500
5,800
5,800

2.12
2.30
2.15
2.73
2.21
2.13
2.18
2.27
2.11
2.19
2.16
2.24
2.07
2.30
2.14
2.19
2.16
2.11
2.05
2.09
2.35
2.05
2.12
2.03
2.05
2.06
2.35
2.35
2.66
2.44
2.06
2.35
2.16

Pile
head
sett

Criteria Not
exceeding(mm)

600Type 2

750

1000

1200

1xWL
2xWL
Residual

12.5
38.0
6.5

5.32
12.85
1.35

4.06
9.39
1.33

5.59
10.67
0.14

4.45
7.37
1.00

WL=working load

Bored pile diameter (mm)6 IMPORTANT OBSERVATION OF STRENGTH OF
PHYLLITE
An important observation made is that despite the very low
RQD obtained for the phyllite rock cores between 10m-18m
depths (actual range of pile socketing depths as installed onsite) and hence very limited UCS test was possible, however,from the back-analysis of the load test results the interpreted
UCS ultimate value of 7.4MPa for phyllite seems to be very
encouraging. It is therefore believed that phyllite, despite
being friable, actually does have a reasonable good in-situ
strength as a mass, which is often neglected because most, if
not all the time, rock coring was carried out and hence RQD is
usually used to interpret the rock strength.
It is believed that the main reason for obtaining a relatively
high UCS of 7.4MPa for phyllite is due to the relatively large
overburden pressure of 10m-18m overlying the rock layer.
This provides a reliable confining stress to the phyllite layer.
The metamorphic phyllite may be well-known to havesmall-scale slip-faults, intense shearing and faulting, but if it
is kept largely under a confining pressure (compression innature) with minimal disturbances, the in-situ strength will
mostly be preserved. For this very reason, if externalmechanical stresses are introduced to a mass of phyllite
during rock coring, phyllite will tend to be disintegrated andwhen the core is retrieved the logging of phyllite as ‘friable’in borehole logsheets is inevitable, hence the injustice of a
misconception that by default phyllite should have a low
strength value.
From this study, the following relationships on phyllite can be
developed based on the successful pile socketing depths ofbetween 10m and 18m on site (considering average values
with respect to their respective standard deviations):170

Unconfined Compressive Strength Test in cored samples
UCS = 7.48 – 11.58 MPa(3)

8 ACKNOWLEDGMENT

Point Load Test in axial direction on cored samples
UCS = k A *I s50(4)where k A = 5.5 – 12.0 and I s50 is the point load index in MPa
Point Load Test using lump samples
UCS = k L *I s50(5)where k L = 0.9 – 6.6 and I s50 is the point load index in MPa
Coincidentally, the preliminary design UCS value of 7.4MPa
seems to coincide with the lower bound value of the UCS test
within the depths of 10m-18m, noting that the I s50 lump
results during design stage was still unknown. The
coefficients k A and k L are applicable to Point Load test on
samples in axial direction and Point Load test on lumpsamples, respectively.
Nonetheless, to increase the confidence level of utilising a
relatively higher UCS design value for phyllite, more data is
to be collected and interpreted in the future. In-situ testing, forexample, pressuremeter test is definitely a recommended
method to obtain a more reliable UCS value for use in design.
The main advantage in using the pressuremeter test is that no
extraction of phyllite sample is required and as such thein-situ strength of the rock mass is preserved and hence can bereliably tested.

The authors would like to acknowledge Jurutera Jasa
(Sarawak) Sdn. Bhd. for their kind co-operation in making
this paper a reality.
9 REFERENCES[1] Ong, D.E.L. and Choo, C.S. Sustainable Bored Pile
Construction in Erratic Phyllite. ASEAN-AustralianEngineering Congress. CD proceeding (ISBN978-967-10485) July 2011.
[2] Tan, D.N.K. Geology of the Kuching Area, West
Sarawak, Malaysia. Report 16, Geological Survey of
Malaysia, Kuching, 1993.
[3] Australian Standard AS 4133.4.1-2007: Methods oftesting rocks for engineering purposes - Rock strength
tests - Determination of point load strength index.
Sydney, NSW, Australia, Standards Australia Limited.
[4] Williams, A.F., Johnston, I.W. & Donald, I.B. The
design of socketed piles in weak rock. Proc. ofInternational Conference On Structural Foundation On
Rock, Sydney, 1980, pp. 327-347.
[5] Ting, W.H. and Fuad, N. Pile Socket Capacity and Rock
Mass Classification. Journal of the Southeast Asian
Geotechnical Society, April 2003.
[6] Basu, A. and Kamran, M. Point load test on schistoserocks and its applicability in predicting uniaxial
compressive strength. International Journal of RockMechanics and Mining Sciences, 47, 2010, pp. 823-828.

7 CONCLUSIONSThis paper has documented the difficulties in quantifying
rock strengths and designing for geological conditions withhigh degrees of fracturing, i.e. RQD of 0%, typical of themetamorphic phyllite of the Tuang Formation. Recovery ofcores for testing is only sporadic at best, requiring the sinking
of more boreholes for the extraction of intact rock cores.
Efforts to correlate UCS with the point load index require
further studies, in particular the lower bound results which
would govern the conservative design of piles socketed inrock.
Further research could realise the great potential in applying
the in-situ pressuremeter test to obtain values of phyllite rockstrength for the design of bored piles. The main advantage inproposing the pressuremeter test is that no extraction ofsample is required and as such the in-situ strength of thestrong mass is preserved and hence reliably tested.
It is important for engineers to be able to interpret themagnitude of phyllite rock strength as this would ultimately
impact the design and costs of bored piles to be installed.

171

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Swelling deformation of the bentonite mixed with silica fume or sodium
carbonate in calcium hydroxide solution
Yuki YOKOYAMA1 and Kenichiro NAKARAI2
Graduate School of Engineering, Gunma University, Japan
2
Graduate School of Engineering, Hiroshima University, Japan
1

ABSTRACT
From these experimental results, it was found that the mixing of small amount of silica fume or sodium carbonate to
bentonite can reduce the deterioration of the bentonite in alkaline water from the viewpoint of swelling performance.
Swelling performance of the bentonite specimen was measured using a one-dimensional swelling-deformation test
under constant vertical pressure .In this paper, we propose a method to solve this problem by the mixing of silica fumeor sodium carbonate. These materials are expected to react with the leached calcium ions from the cement-basedmaterials and reduce the deterioration of the bentonite. It decreases swelling performance of the bentonite. However,during the tens of thousands of years, bentonite can be deteriorated and dissolved by alkaline water with the leachedions from the surrounding cement-based materials. In radioactive waste disposal facility, sodium-type bentonite will be
used as an artificial barrier material.
Keywords: artificial barrier, bentonite , silica fume , sodium carbonate , swelling deformation
1. INTRODUCTION
In radioactive waste disposal facilities, sodium bentonite
having high swelling capacity will be used as an artificial
barrier material. It is expected to delay the infiltration ofgroundwater and to fill cracks of the surrounding host rock.
However, during tens of thousands of years, the bentonite canbe deteriorated by the alkaline water with the leached ionsfrom the surrounding cement-based materials. It decreases
swelling capacity of the bentonite [1]. To reduce the decreasein the swelling capacity by the alkaline water, the authors
have proposed two kinds of admixtures. One is the mixing of
pozzolanic materials such as silica fume and fly ash to thesodium bentonite [2]. These pozzolanic materials are
expected to consume the leached calcium ions from thecement-based materials and to reduce the deterioration of thebentonite. Experimental results in the previous research [2]showed that the mixing of the pozzolanic materials can
reduce the decrease in the swelling capacity. In theexperiment, the one-dimensional vertical swellingdeformation was measured by using the swelling-deformation
tests proposed by Komine and Ogata [4], [5]. Especially, themixing of 1 mass% of silica fume was the most effective to
reduce the deterioration by calcium hydroxide solution. But,
the detailed investigation should be added because the rangeof experimental conditions was limited.
The other proposal is the mixing of sodium carbonate to thebentonite. Here, calcite precipitation between the
bentonite-based materials and cement-based materials was
expected. The previous experimental results conducted by
Nakarai et al [3] showed the calcite precipitation can reducethe deterioration of the bentonite due to calcium leachingfrom the cement paste. Here, the immersion tests of thecomposite specimens consisting of the cement paste and thebentonite were performed. Then, the effect of the mixing of
the sodium carbonate was investigated from the distributionof calcium measured by Electron Probe Micro Analysis(EPMA). But, the swelling characteristics of the sodium
carbonate mixed bentonite haven’t been studied.
In this study, we investigated the effect of the mixing ofsilica fume or sodium carbonate to bentonite on the swellingcharacteristics of the bentonite. To investigate the swellingcharacteristics, we also used swelling-deformation tests
proposed by Komine and Ogata [4], [5]. Then, we measured
swelling deformation of the compacted bentonites whichwere immersed in the distilled water or the saturated aqueous
solution of calcium hydroxide. As preliminary experiments,
we also studied the influence of the experimental condition
such as means of water supply and concentration of calcium
hydroxide solution. Then, we decided the condition toappropriately measure the swelling deformation of thebentonite specimens in this study.
2. OUTLINE OF EXPERIMENTS
2.1 Materials for specimens
Table.1 shows the outline of specimens. In this study, weused silica fume or sodium carbonate as an admixture. Then,
we prepared three kinds of specimens; pure bentonite without
admixtures, bentonite mixed with silica fume, and bentonite
mixed with sodium carbonate. This study used commercial
bentonite, called Kunigel-V1, extracted from the Tsukinuno
Mine in Japan. This study also used silica fume. The 94.9% ofsilica fume was SiO 2 and its specific surface area was 17.5
m2/g. This study also used commercial sodium carbonate. Its
content of sodium carbonate was 99.8%. The mixing ratios ofsilica fume to dry bentonite were 0.5, 1.0, 2.0, 5.0, 8.0 mass%
while the mixing ratios of sodium carbonate were 1.0, 4.0
mass%. The water content of bentonite was set to 21 % byconsidering the planned value in the design of a subsurface

172radioactive waste repository for low level wastes in Japan. In
this paper, the names of specimens indicated the compositionof the materials, the mixing ratio of admixture, the type ofsolution (W: distilled water, CHs: saturated calcium
hydroxide solution, CHp: saturated calcium hydroxide
solution with calcium hydroxide powder) and supply ofsolution (1: one side, 2: two sides). The supplying method
will be discussed only in chapter 3.

Table1．1 Outline of specimen
Names ofspecimens
Be-W
BeSF-0.5-W
BeSF-2-W
BeSF-3-W
BeSF-5-W
BeSF-8-W

2.2 Mixing of materials
The mixing process is as follows. First, air-dried bentonite
and admixture were mixed at low speed for 30 seconds using
an Omni mixer. During the mixing, water contents of thesemixtures were also adjusted. Here, we sprayed distilled water.
After that, they were mixed at high-speed for 60 seconds. And
they were kept in a polyethylene bag for 24 hours. Next day,these mixtures were mixed at high-speed for 60 seconds
again, and we measured water contents of these mixtures.
2.3 Specimens
In this study, cylindrical specimens having diameters of
60mm and heights of 5mm were prepared. The height of thespecimens became small to save time for water absorption
because the compacted bentonite specimens have low
conductivities. The dry density was set to 1.3g/cm3 to easily
assess deterioration. The sample was statically compacted byvertical pressure for several hours until the height ofspecimens became 5mm.
2.4 Swelling-deformation test
Fig. 2.1 shows the test apparatus. Swelling characteristics ofthe bentonite specimens were evaluated by using the
one-dimensional swelling-deformation test under a constantvertical pressure proposed by Komine and Ogata [4], [5]. Iron
plate was fixed to the top of the piston applying the constantvertical pressure. We used a filter paper, of which thediameter is 60mm and the thickness is 0.19mm, and a porous
metal, with 60mm diameter and 5mm thickness, as filter. The
filters were placed below the top cap and on the pedestal.
They were used for mitigating the outflow of bentonite
particles from specimens.
After applying the prescribed vertical pressure (10kPa) tothe specimen, aqueous solutions were supplied to thespecimen. Then, the relationship between the axial swellingdeformation and the time required from the start of watersupply was measured.
The solutions supplied to the specimens were three kinds ofaqueous solutions. The first solution was the distilled water tomeasure basic performance. The second and third solutions
were the saturated aqueous solutions of calcium hydroxide to
discuss the effects of calcium ions leached from
cement-based materials. The third solution included
additional calcium hydroxide powders to keep saturated
concentration, while the second solution without additional
calcium hydroxide allowed decrease in the concentration
during experiments.
In the cases of specimens of the pure bentonite without
admixtures and the bentonite mixed with silica fume, thetemperature during the experiments was set to 40 °C in order

Material composition[mass%]

Type of Supply ofsolution solution

Be

SF

Na2CO3

100.0
99.5
98.0
97.0
95.0
92.0

0.0
0.5
2.0
3.0
5.0
8.0

0.0
0.0
0.0
0.0
0.0
0.0

distilledwater

saturated one side
calciumorhydroxide
two
solutionsides

Be-CHs

100.0

0.0

0.0

Be-CHp

100.0

0.0

0.0

BeSF-0.5-CHp

99.5

0.5

0.0

BeSF-1-CHp

99.0

1.0

0.0

BeSF-2-CHp

98.0

2.0

0.0

BeSF-5-CHp

95.0

5.0

0.0

BeSF-8-CHp

92.0

8.0

0.0

BeNa2CO3-1-CHp

99.0

0.0

1.0

BeNa2CO3-4-CHp

96.0

0.0

4.0

one sidesaturatedcalciumhydroxide
solution
one sidewithcalciumhydroxidepowderLinearVariable
displacementtransducer
Iron plate
PistonSpecimen
Diameter:60mm
Height:5mm

Acrylic cell
Solution

Pedestal

Porous metal

Fig．2.1 Schematic drawing of test apparaｔus

to promote pozzolanic reaction of silica fume. In the case of
specimens of the bentonite mixed with sodium carbonate, thetemperature was set to 20 °C. They were determined byconsidering the experimental condition in the previous
studies [2], [3].
The solution was supplied from one side of specimens in thisstudy. Although the water supply from two sides could savetime for measurements, the condition in this study wasdecided to be same as the previous study [2] to easily compare
the results. The validity of this water supplying condition wasinvestigated by the preliminary experiments.
Measurement period was also investigated by the
preliminary experiments. Then, it was decided to be 7 days. It
was longer than the previous study conducted by Nakarai and
Shimakura [2] to improve the accuracy of data.

173

2.5 Evaluation method
The swelling strain (%) is defined by
∆S
× 100
H0

Swelling strain εs[%]

εS ＝

Be-CHs-1(1)

where εS is swelling strain (%), ∆S is swelling deformation,and H0 is initial specimen height.
It is known that the curves of swelling strain versus time canbe approximated by the hyperbola in Eq.(2) [4], [5]. The
maximum swelling strain εsmax (%) is obtained from theasymptotic line of the hyperbola, and can be calculated by
Eq.(3)εs (time)＝

time
[%]
a + btime1
εsmax = lim εs (time) = [%]time→∞b(2)(3)

where “time” is the duration (in day) from the start of watersupply, ε s (time) is swelling strain at “time” , and a and b are
constants determined by fitting procedures.
3 RESULTS OF PRELIMINARY EXPERIMENTThis chapter describes the results of the preliminary
experiment to decide the experimental conditions of theswelling-deformation test.
3.1 Influence of additional calcium hydroxide powder
Fig.3.1 shows the results of the swelling-deformation tests
investigating the influence of additional calcium hydroxide
powder. They show the relation between the curve ofswelling strain versus time and the calculated maximum
swelling strain. The swelling strain of Be-CHs-1 was greaterthan that of Be-CHp-1. We supposed that the reduction of thedeterioration of the bentonite in the case of Be-CHs-1 without
additional calcium hydroxide powders was caused by thedecrease in the concentration of calcium ions in the solution.
Deterioration of the bentonite in the case of Be-CHp-1
including additional calcium hydroxide powders is
progressed because the solution kept saturated concentration.
To keep saturated concentration is necessary to correctly
evaluate swelling performance of the bentonite.
Therefore, we used the solution included additional calcium
hydroxide powders in the experiment described in the next
chapter.
3.2 Influence of supplying method of solution
We studied influence of the supply of the solution toswelling strain of specimen. Fig.3.1 shows the results of theswelling-deformation tests on the influence of the supply ofsolution. The swelling deformation rate of Be-CHs-2 wasrelatively fast and the swelling was almost converged in threedays. On the other hand, the swelling deformation rate of
Be-CHs-1 was relatively slow and the swelling deformation
was gradually increased until 7 days (the end of
measurement). In the specimen of Be-CHs-2, the rate of the
supply was faster than that in Be-CHs-1 because the saturated

250
200
150
100
50
0

Be-CHs-2

Be-CHp-1

Maximum swelling ratio
εsmax [%]
ε
[%]
smax

0

2

4

6
Time[day]∞Fig.3.1 Curves of swelling strain versus time
and the maximum swelling strain

calcium hydroxide solution was supplied from the bottom andtop of the specimen. This increase in the solution supply ratecaused the increase in the swelling deformation. The resultsof the maximum swelling strain of Be-CHs-1 and Be-CHs-2
are approximately same. It means that the supplying methodof the solution doesn’t affect the maximum swelling strain.
4 RESULTS AND DISCUSSIONS
Fig.3.2 shows the results of the swelling-deformation tests
investigating the swelling deformation when the distilledwater (W) was supplied to the specimens which were mixed
with silica fume. Fig.3.3 shows the results of theswelling-deformation tests when the saturated calcium
hydroxide soluteon with additional powders (CHp) was
supplied to specimens which were mixed with silica fume.
Fig.3.4 shows the relationship between the calculated
maximum swelling strain and the mixing ratio of silica fume.
Here, the solid and broken lines in the figure show thepredicted maximum swelling strain in W and CHp,
respectively. They were predicted from the maximum
swelling strain of the pure bentonite specimens and themixing ratio. Fig.3.5 shows the change in the measured
swelling strain with time and the calculated maximum
swelling strain.
4.1 Effect of silica fume as an admixture
According to Fig.3.4, in the case of W, the maximum
swelling strain of specimen was smaller than the predictedmaximum swelling strain indicated by the solid line.
Therefore, it was found that the maximum swelling strain was
decreased by the mixing of silica fume when W was supplied.
We suppose that it was caused by solidification with the
pozzolanic reaction of silica fume in the water. In the case ofCHp, it was found that the maximum swelling strain of
specimens mixed with silica fume (0.5 ~ 5 mass %) became
larger than the predicted maximum swelling strain indicatedby the solid line. It means that the mixing of silica fume canreduce the deterioration of bentonite by alkali. Especially, themixing of 1mass% of silica fume was most effective tosuppress the deterioration. When we mixed silica fume morethan 1mass%, the maximum swelling strain gently decreased
as mixing ratio of silica fume increase. It was caused by thedecrease in the bentonite proportion in the specimen and

174

0

2

BeSF-0.5-CHp
BeSF-1-CHp
BeSF-2-CHp

BeSF-2-W
BeSF-3-W
Swelling strain εs[%]

Swelling strain εs[%]

250
200
150
100
50
0

BeSF-0.5-W
BeSF-8-W
BeSF-5-W

4

6

8

250
200
150
100
50
0

0

2

Fig.3.4

Fig.3.3 Curves of swelling strain versus time

BeSF-CH
Swelling strain εs[%]Maximum swelling strain ε smax[%]

250
200
150
100
50
0

BeSF-W

The relationship content ratio
of bentonite in the specimen
and maximum swelling strain

0

2

4

6

8

Time[day]Fig.3.2 Curves of swelling strain versus time

Be-CH

6

4

Time[day]

Be-W

BeSF-3-CHp
BeSF-5-CHp
BeSF-8-CHp

8

10

250
200
150
100
50
0

BeNa2CO3-1-W

BeNa2CO3-1-CH

BeNa2CO3-4-W

BeNa2CO3-4-CH
εsmax [%]

0

2

4

6
Time[day]Mixing rate of silicafume [mass%]∞

Relationship between maximum swelling strain
and mixing rate of silica fumeFig.3.5 Relationship between maximum swelling strain
and curves of swelling strain versus timesolidification associated with the pozzolanic reaction of silica
fume in the water.

(1) The mixing of small amount of silica fume can reduce thedeterioration of the bentonite in alkaline water.
(2) The mixing of 1 mass% of sodium carbonate can reducethe deterioration of the bentonite in alkaline water.

4.2 Effect of sodium carbonate as an admixture
According to Fig.3.5, in the case of the mixing of 4 mass%of sodium carbonate, the specimens both in W and CHp
showed similar swelling behaviors. It means that the mixingof 4 mass% of sodium carbonate did not reduce thedegradation of bentonite. In the case that mixing ratio of
sodium carbonate was 1mass%, there were no differences inthe swelling behaviors until 2 days. But, after 2days, it was
found that the swelling strain of BeNa 2 CO 3 -1-CHp was
greater than that of BeNa 2 CO 3 -1-W. It was supposed to becaused by the decrease in the ionic strength of the pore waterin the particles associated with the precipitation of calcite.
5 CONCLUSIONSIn this study, we investigated the effect of admixtures suchas silica fume and sodium carbonate in order to develop newmaterials suppressing the deterioration by alkali from theviewpoint of swelling performance. The performance was
evaluated based on the swelling deformation of specimens
measured in the distilled water and the saturated calcium
hydroxide solution. Then, the following conclusions weredrawn from the results of the experiments.

6 REFERENCES[1] Japan Society of Civil Engineers, “The method for setting
nuclide migration assessment parameter used in groundwater
migration scenario in safety assessment of sub-surface
disposal,” 2008. (in Japanese)[2] Nakarai K., and Shimakura C., “Swelling deformation of
the bentonite mixed with pozzolanic materials in calcium
hydroxide solution,” Proceedings of 46th Conference of theJapanese Geotechnical Society, 2011, pp.2089-2090. (inJapanese)[3] K. Nakarai, M. Watanabe, H. Ishii, K. Koibuchi: Reducethe deterioration between cement and bentonite mixed withNa 2 CO 3 , Proceedings of the 66th Annual Conference of theJapan Society of Civil Engineers, pp.99-100, 2011.
[4]H. Komine, H. Ogata: Experimental study on swellingcharacteristics of compacted bentonaite, Canadian
Geotechnical Journal, Vol.31, No.4, pp.478-490, 1994.
[5]H. Komine, H. Ogata: Experimental study on swellingcharacteristics of sand-bentonite mixture for nuclear waste
disposal, Soils and Foundations, Vol.39, No.2, pp.83-97,
1999.

175

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

An Interpretation of Mechanical Properties of Bentonite as a Non-linear
Elastic Material
Y.TAKAYAMA, S.TSURUMI, A.IIZUKA, K.KAWAI
KOBE UNIVERSITY, JAPANS.OHNOKAJIMA CORPORATION, JAPANABSTRACTDue to its low permeability and remarkable expansion characteristics, bentonite is an excellent candidate with potential
use as a buffer in the disposal of nuclear waste. In order to examine long term safety of bentonite buffer in the disposalof nuclear waste, constitutive modeling of bentonite materials and numerical simulations into which its constitutive
modeling is incorporated have been actively developed in recent years. The bentonite materials are usually modeled asa geo-material composed by elasto-plastic isotropic compression/expansion and plastic dilatancy characteristics.
However, based on recent experimental data, the authors have begun to think that such usual treatment of constitutive
modeling for geo-materials may not be applicable to the bentonite materials. In this paper, fundamental considerations
in elasto-plastic constitutive modeling for fully saturated bentonite materials are made. According to uniaxial
compression test data by Sasakura et al (2003) in which lateral earth pressure was measured, it is found that thehysteresis response between loading and unloading processes does not appear. It suggests that the bentonite is elastic
materials in which the swelling line corresponds with the normally consolidation line in the e-logp’ relationship. Also,
according to triaxial CU test results of bentonite materials, it can be seen that the effective mean stress is almost
unchanged during shearing. This implies that the bentonite does not have dilatancy characteristics based on the criticalstate theory. Similar interpretation can be derived from the experimental results obtained from a series of triaxial CD
tests by Cui et al (2006).
1. INTRODUCTON
Bentonite is known to be a material that has remarkable
expansion characteristics and very low permeability.
Therefore, the bentonite material is expected to use as a buffer
in the geological disposal of nuclear waste facility and its
mechanical and mineralogical properties have been
investigated. In the nuclear disposal site, the role to prevent
leaking of radioactive polluted water by self-seal ability andperformance to mitigate stresses generated due to creep ofbase rock are expected for the bentonite as a buffer. Also, thestability and safety for an extremely long period of time are
required for the geological disposal of nuclear facility. In
order to address such problems, it is necessary to predict thelong term behavior of the geological disposal of nuclear waste
facility and reliable numerical simulation technique thatenables it is required The constitutive modeling of a material
which describes the mechanical properties is an essential key
technique in the numerical simulation. As for the constitutive
modeling of bentonite materials, it has been formulated
phenomenologically based on the elasto-plastic response
under the loading and unloading processes and dilatancy
characteristics during shearing like other clayey materials.
However, in case of bentonite materials, the authors think that
explicit experimental supports to such constitutive modelingare not sufficient. Remarkable expansion characteristics onwater absorption and very low permeability of the bentonite
make it difficult to interpret their fundamental mechanicalbehavior. The aim of this paper is to reconsider thefundamental mechanical behavior of the bentonite materials

based on recent experimental data and to discuss thetheoretical formulation of constitutive modeling for the
bentonite materials.
2. Isotropic compression and swelling characteristics ofsaturated bentonite materials
Usually, isotropic compression and swelling characteristics
are discussed based on the experimental data obtained fromtriaxial tests under the drained condition. However, in thecase of bentonite materials, there is not sufficient reliable dataon the compression and swelling characteristics obtained
from triaxial tests under the drained condition.
Manyoedometer test data have been reported instead.
Compression and swelling behavior of the bentonite materials
obtained from the oedometer tests is not directly available indiscussing compression and swelling characteristics underisotropic confining pressure, because the lateral stress cannot
be measured in the oedometer tests. If the lateral stress can bereliably estimated, more intensive discussion and more strict
interpretation on the compression and swelling characteristicsof the bentonite materials is thought to be made possible. In
this paper, the authors cite the experimental data of measuring
the lateral stress with a special oedometer test apparatus forthe bentonite materials by Sasakura et al. (2003) and discuss
the compression and swelling characteristics of the bentonite
materials under isotropic confining pressure, which are
fundamental mechanical properties for constitutive modeling.

176

m of 0.6 in converting the vertical stress into the isotropic
confining pressure, fundamental mechanical properties oncompression and swelling characteristics under the isotropic
confining pressure can be discussed using experimental dataobtained from the oedometer test.

10
5

1
0.5
1

in which K 0 − NC : the coefficient of earth pressure at rest under
the normally compression process, K 0 − OC : the coefficient ofearth pressure at rest under the unloaded swelling process,
OCR : over-consolidation ratio defined by σ v′0 σ v′ , σ v′0 :pre-consolidation vertical stress, σ v′ ：current vertical stress,

m ：evaluation constant.

According to Fig.2, Eq.(1) with

K0

1
0.9
0.8
0.7
0.6
0.5
4

Fig.2

5

10
OCR

50

100

The relationship between Ko value and OCR（2）Isotropic deformation characteristics
Fig.3 shows one dimensional compression (consolidation)
characteristics of Kunigel V1 obtained from the oedometer
tests by Sasakura et al. (2002) and Ishikawa et al. (1997).
Normally consolidation lines by Sasakura et al.(2002) andIshikawa et al.(1997) does not correspond each other. This
would be because the nature of the bentonite slightly differs
depending on the year when the bentonite is produced and thecontent of the montmorillonite is slightly different eventhough the same bentonite.
According to Kobayashi et al. (2007), one dimensional
normally consolidation characteristics of the bentonite under
0.8

3Ishikawa et. al ,1997
Ishikawa et. al ,1997
Sasakura et. al ,2002
Fully saturation curve

2
1

Void ratio, e

Horizontal effective stress,
σ'r MPa

Void ratio, e

2

m=0.4
m=0.5
m=0.6
Test result

K0（1）K 0 compression and swelling test
Sasakura et al. (2003) carried out one dimensional
compression and swelling tests for Kunigel V-1, which is atypical material of sodium bentonite and is planned to be
used in the nuclear waste disposal site, Japan. And they tried
to measure the lateral stress under one dimensional loading
and unloading processes with a special oedometer test
apparatus. Their experimental results are shown in Fig.1.
It is found that the coefficient of earth pressure at rest, Ko, isalmost kept 1.0 under the normally compression process andthen increases with OCR under the unloaded swelling
process. The relationship between Ko value and OCR isindicated in Fig.2. In this figure, experimental data by
Sasakura et al. (2003) are indicated by white circles and thefollowing empirical relation, Eq.(1), between Ko-value andOCR with the evaluation parameter, m, is compared,
K 0 − NC = K 0 − OC OCR m （1）

10

0.6

4
0.4

2
1

0.01
0.1
1
10
100
Vertical effective stress, σ'v (MPa)0.010.050.1
0.5 1
Vertical effective stress, σ'v (MPa)
Fig.1

Fig.3 One dimensional compression (consolidation)
characteristics of Kunigel V1

K0 compression and swelling test result

m of 0.6 seems to well explain experimental results of thebentonite, Kunigel V-1. If it is allowed to employ Eq.(1) withfully saturated state is uniquely expressed as,

177

e

λρ ln

σ v′ maxσ v′

σ v′ max  Gs − λρ ln ′ σv σ v′ max = 3.73GPa .

（2）

Then, the effective vertical stress can be converted into theeffective mean stress using Eq.(1) as,
σ ′ + 2 K 0σ v′（3）p′ = vin which λρ : material constant, Gs : specific gravity of soil
particle, σ v′ max : reference and maximum vertical stresscorresponding to the void ratio vanishing. This curve iscalled as fully saturation curve. In the case of data by
Sasakura et al. (2002), λ ρ = 0.156 , G s = 2.733 ,Void ratio, e

0.7

Sasakura et.al, 2002
Fully saturation curve

0.6

0.5
0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)ｍ＝0.6 are employed. It is found that the hysteresis loop
between loading and unloading processes disappears and theswelling line corresponds with the compression line in Fig.4.
If the compression and swelling responses in Fig.4 can beinterpreted as isotropic compression and swellingcharacteristics of the bentonite, the bentonite can be regarded
as a material having no plastic deformation. Namely it can besaid that the bentonite is a non-linear elastic material.

0.4

3. Dilatancy characteristics of saturated bentonite
materials
In this section, we examine the dilatancy
characteristics of bentonite. Usually, dilatancy
characteristics are examined conducting triaxial CD
test in which the principal mean effective stress ismaintained constant. However, in the case of
bentonite materials, there is not sufficient reliabledata obtained from this test. Therefore, we examine
dilatancy characteristic using the results of triaxial
CU test on bentonite.

0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)（1）Triaxial CU test
Results of triaxial CU test on some kind of bentonite
are shown in Fig.5-8. These figures show the
relationship between effective mean stress and

(A)

0.6

Void ratio, e

3

in which p ′ : effective mean stress, σ v′ : effective verticalstress, K 0 : coefficient of earth pressure at rest. All data in
Fig.3 are converted in terms of effective mean stress using
Eq.(3) and are depicted in Fig.4 as a relation between the void
ratio and the effective mean stress, in which K 0 − NC ＝１ andIshikawa et. al, 1997
Fully saturation curve

Deviator stress , q (MPa)(B)

Void ratio, e0.8
Ishikawa et. al, 1997
Fully saturation curve

0.6
0.4
0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)

3
2
1

0
0 1 2 3 4 5 6
Effective mean stress ,p' (MPa)(C)
Fig.4 Isotropic compression and swelling characteristics ofsaturated bentonite materials

Fig.5 Triaxial CU test result on Sodium type of
Bentonite conducted by Sasakura et al. (2002)
178

Deviator stress , q (MPa)

Deviator stress , q (MPa)

5
4
3
2
1

0
0 1 2 3 4 5 6
Effective mean stress ,p' (MPa)

Deviator stress , q MPa

4
3
2
1

0
0
5
10
Effective mean stress ,p' (MPa)Fig.6 Triaxial CU test result on Calcium type of
Bentonite conducted by Sasakura et al. (2002,2003)

Fig.8 Triaxial CU test result on MX-80 conductedby Ann et al. (2010)
under uniforming the volume. Typically, these result isarranged as connection between pressure at saturation to dry
density. However, the relationship between these test’s
results and full saturation line is not elucidated perfectly.
Saturated bentonite is the elastic body, so its condition isalways on full saturation line. Because of this, we estimatethat the final value of these swelling test get to full saturation
line. Now, we examine relationship between swellingcharacteristic and full saturation line, based on the result of
these swelling test.

2

1

0
0
1
2
3
Effective mean stress ,p' (MPa)

Fig.7

5

Triaxial CU test result on OT-9607
conducted by Takaji et al. (1999)

deviator stress. According to these test results, it canbe seen that the effective mean stress is almost
unchanged during shearing. Usually, the effective
mean stress decreases with increasing shear stressbecause of negative dilatancy when triaxial CU test
on normally consolidated clay is conducted.
Therefore, it is implied that the bentonite does nothave dilatancy characteristics based on the criticalstate theory.
4. Swelling characteristic of saturated Bentonite
As one of important mechanical properties of bentonite,
there are swelling characteristic that is found in process fromunsaturated to saturated. We have two tests, swellingdeformation test and swelling pressure test, as a way ofmeasuring of swelling characteristic. Swelling deformation
test teaches us deformation when the bentonite suck the water
under controlling the stress, and the swelling pressure test
teach us swelling pressure when the bentonite suck the water(1)bentonite’s swelling deformation test
Cui(2006) conducted swelling deformation test andisotropic unloading test with controlling principal stress. This
test was performed on the sample of compacted Kunigel
V-1/Toyourasa mixture with Kunigel V-1 content of 50% in
dry mass. In this test, they set the initial dry density threepatterns,1.69,1.54,1.2g/m3,and then did water absorption testwith hydrostatic stress, based on the condition that the test
keep mean principal stress settled. We show the sample’s
initial value, initial dry density and initial water content ratio,
and the value of mean effective principal stress which thesample suck the water on the table-1.
In addition, they did isotropic unloading test which reduce theload under 49kPa after the sample suck the water with mean
principal stress set 392kPa. Fig-9 show these result,relationship between mean effective principal stress and void
ratio at the finish of retting and swelling, of these three case of
initial dry density and relationship between mean principal
stress and void ratio when the test goes on isotropic
unloading. In the case that it gave the sample some water insame mean stress, Fig-9 shows that the void ratio takes samevalue regardless of initial dry density. Furthermore, the valueof unloading pass at isotropic unloading shows the almost
same result with the swelling deformation test. Now, we takethe three words, full saturated line/normal consolidation
line/isotropic unloading line, as the same meaning. Based onthese result, we can find that the value in case of swelling testis on full saturated line at saturated condition. Next, they
conducted swelling test with the different stress ratio179

R( = σ a / σ r ) in which the principal mean stress and principal
stress ratio are maintained constant. The initial value of thesample takes 1.69g/m3 as dry density and 18.7% as initialwater content ratio. We show the relationship between void
ratio and mean principal stress at the end of soaking in Fig-10.
Fig-10 teaches us that void ratio and mean stress are
relationship of one to one regardless of principal stress ratio.
So, our examination that saturated bentonite is nonlinearly
elastic material without characteristic of dilatancy coincides
with the result of Cui’s test.
ρd0（g/cm3）
1.69
1.54
1.2

10
5

Void ratio , eρd0=1.69g/cm3
ρd0=1.54g/cm3
ρd0=1.20g/cm3
Isotropic unloading
392→49kPa

1
0.5

50 100500 1000

Effective mean stress , p' (kPa)void ratio , e

Fig.9

the result of amount of swelling test

10
5

R=2.0
R=1.5
R=1.0

50

100

500

Effective mean stress ,p' (kPa)
Fig.100.51
5 10
50 100Swelling pressure (MPa)the result of swelling pressure test(2)Swelling pressure test of bentonite
Swelling pressure test, using the equipment similar to
one-dimensional consolidation equipment, is the test which
can measure counter force of vertical direction caused byvertical soaking under settled volume. Fig-11 shows the resultof swelling pressure test of Kunigel V-1, and shows
relationship between void ratio and vertical swelling pressure
at saturated. This figure teaches us that void ratio is smaller,
then swelling pressure is larger and swelling pressure takes
apart value in the same void ratio, but the value doesn’t
exceed the full saturated line. But, considering the result so
far obtained, these apart value results from measuring only
vertical swelling pressure. So, if we take account of lateralstress and reconsider by the use of mean stress, and we
consider that the value is all on the full saturated line.
5. Conclusion
This study examines the mechanical characteristics ofbentonite based on recent experimental data. The conclusions
are summarized as follows.
(1) The swelling line at S r = 1.0 corresponds with theNormal Consolidation line. And saturated
bentonite does not have dilatancy characteristics.
Therefore, saturated bentonite is a non-linear
elastic material.
(2) When water is supplied from unsaturated state
to saturated state, the final or saturation points
reach on the fully saturation curve.

1
0.5
0.1
10

Void ratio, e

Fig.11

18.7
49,98,147,196
18.7 19.6,49,88,196,294,392
19
49,98,196,392
Table-1 initial value

Fully saturation curve
Suzuki et.al, 1999
Komine et.al, 2002

0.5
0.1

p
（kPa）

w0
(%)

0.1
10

1

In Fig. 12, the schematics of relationship between void ratio
and the principal mean stress during loading, unloading andswelling are summarized.

the result of swelling test180

eFully saturation curveor
NCL or Swelling line
at Sr=1.0

InitialSwelling
Pressure Test

Swelling
Deformation
Test

Initialln pFig. 12 Schematics of relationship between void ratio
and the principal mean stress

REFERENCES[1] SASAKURA,T., KUROYANAGI,M.., KOBAYSHI,I.,
OKAMOTO,M. ”Studies on mechanical behavior of
bentonite for development of the constitutive model II”,
2003 ,JNC TJ8400 2003-048. (in Japanese)[2] SASAKURA,T., KUROYANAGI,M., OKAMOTO,M.,“Studies on mechanical behavior of bentonite fordevelopment of the constitutive model”, 2002, JNC
TJ8400-2002-025. (in Japanese)[3] ISHIKAWA,H.,
ISHIGURO,K.,
NAMIKAWA,T.,
SUGANO,T., “Consolidation Properties of Buffer
Material” 1997, PNC TN8410 97-051 . (in Japanese)[4] KOBAYSHI,I., TOIDA,M., SASAKURA,T. and
OHTA,H., “Interpretation of compression/swelling
behavior of compacted bentonite using
constant
water-content line and constant degree-ofsaturation line”,
2007,JSCE Journal of GeotechnicalEngineering,
63(4):1065-1078. (in Japanese)[5] TAKAJI,K., SUZUKI,H. ”Static Mechanical Properties
of Buffer Material” 1999, JNC TN8400 99-041 (inJapanese)[6] Ann Dueck, Lennart Börgesson, Lars-Erik Johannesson,“Stress-strain relation of bentonite at undrained shear”
2010, SKB Technical Report TR-10-32[7] Hongbin CUI, De’an SUN, MATSUOKA,H., “Swellingcharacteristics of Sand-Bentonite mixtures underisotropic and anisotropic stress States” 2006, JSCE
Journal of Geotechnical Engineering, 62(3): 657-666.
(in Japanese)[8] SUZUKI,H. and FUJITA,T, “Swelling Characteristics of
Buffer Material” 1999, JNC TN8400 99-038. (inJapanese)[9] KOMINE,H. and OGATA,N., “Swelling characteristics
of sand-bentonite mixture and various kinds ofbentonite”, 2002, JSCE Journal of GeotechnicalEngineering, 701(III-58):373-385. (in Japanese)

181

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Parametric Study of Crossing Tunnels at Different Level
A. Tohidi, MH.Sadagiani
School of Civil Engineering, Sharif University of Technology, Iran,
ABSTRACT
Tunnel construction in congested urban areas is an essential part of transportation system development. In this study aparametric Analysis was conduct to study the effect of tunneling on the existing support system (i.e. shotcrete lining) ofa shallow tunnel and Surface Settlement. The parameters which can be studied are Soil characteristics, construction
procedure and angle of intersection. Finite element method with Mohr-coloumb constitutive model was applied fornumerical simulation of excavation process.
Parametric analysis shows notable importance for each parameter and we are able to minimize the Surface Settlementwith some suitable considerations.
boring of shallow tunnel firstly, soil improvement in concerned area and selection of a perpendicular angle inintersection region are some instances that will decrease Surface Settlement.
Keywords: Crossing tunnels at different level, Numerical modeling, parametric analysis

represent soil and tunnel lining behavior are mohr-coloumb

1. Introduction
There are different applications for underground structures inmetropolitan area including transportation, lifeline utilities
and etc. These various applications show the importance andnecessity of using underground spaces. Underground metro
lines and underground facilities are a suitable solution for
traffic problems; therefore, probability of locating a newtunnelnearbyotherexisting

undergroundstructures

incredibly increases. In these cases, the influence of newtunnel construction on increasing the ground surface
settlement and bending moment of existing structure should
be investigated. Generally, there are four approaches forpredicting the surface settlement and bending moment indifferent

steps

of

tunneling,includinglaboratory,

experimental, analytical and numerical methods. In thisarticle, 3D numerical simulation has been applied by Abaqusand linear elastic, respectively.
The effect of Important parameters such as construction
procedure,intersection angle and soil improvement are
Studied. In conclusion, it has been shown that the proper
selection of these parameters minimizes the effect of newtunnel construction on surface settlement or bending momentof existing tunnel.
As before mentioned there are three main methods to study
the interaction of nearby tunnels containing: field observation[1], physical model tests [2] and numerical simulation
methods [3, 4, 5]. Liu (2009) presented a good summary
about studies conducted until know.
In general Most of the numerical modeling is 2D and in 3Dmodel a brief study about the influence of parameter likeangle of intersection.

finite element software. The constitutive models which

182

2. Geometry and mesh of the modelof excavation stages in base model are 53 stages, in stage 1

The dimensions of whole model according to figure 2-1 isthe gravity load applied and stage 2 to 27 is related to111 x 111 x 70m, also tunnels with polycentric section haveexcavation of shallow tunnel and from stage 28 to 53 deep

been considered according to figure 2-2. The depth of shallow

tunnel excavated. For the stability of analysis the 6m at thetunnel is 10 m and the depth of deep tunnel is 22 m the

begin and end of the tunnel weren’t excavated.

distance between the floor of shallow tunnel and crest of deepThis technique caused more reasonable result than when thetunnel is 3.5 m. the properties of soil and initial lining haveexcavation starts at the boundary condition.

been shown in table I. try and error in modeling, showed that

Unsupported span length is 1.5 m at 12 m before and after thefor tunnel area and around it, the use of Wedge elements isintersection point and 3m at the rest.

more convenient than Structure(cube)elements, so the Wedge

In one stage the soil element is inactive and in next stage theelements have been used around the shallow and deep tunnels,

lining element will be active.

and the eight-node isoparametric hexahedral elements with

All parameters that study are presented in table II.

reduced integration (C3D8R) used in rest cases.

3. Results
For gaining the influence of interaction, we must study at firstwhat happen in boring the shallow tunnel and deep tunnel
alone and then compare the result when both of them
excavated. Thus in figure 3.1, the surface settlement ofshallow tunnel alone and both tunnels have been shown andin figure 3.2, the surface settlement of deep tunnel alone andboth tunnels have been shown. In figure 3.3, the diagram ofbending moment in shallow tunnel crest during the shallow
tunnel and deep tunnel excavation have been shown.
Table I- specifications of base modelFig 2-1.The Geometry, Mesh and Monitored Paths

Parameters
Internal friction
Angle
Cohesion(kPa)
Poissoncoefficient
Elastic module(MPa)
Density(kg/m^3)

Fig 2-2. section of excavated tunnel
27480 elements used in soil area and 7104 element used for
initial lining. In simulation of lining, the four-node shellelements with reduced integration (S4R) have been used. The
way of mesh modeling is such that whatever we go away fromtunnels, the mesh density proceeds to attenuate. The numbers
183

Soil:
Mohr-Colomb- K 0 =0.5

Initial lining:
30 cm- linear elastic

30o

-

50

-

0.3

0.15

50

30000

2000

2400

Table II – parameters that study
Descriptions

Parameters

1. first the shallow tunnel then deep tunnel excavatedConstruction Procedure

2. first the deep tunnel then shallow Tunnel excavated
3.The shallow and deep tunnels excavated simultaneously

Soil Characteristics

Cohesion(Kpa)

Friction
Angle

One

65

40o

100

Two

55

35o

75

Three
(base)

50

30o

50

100

30o

200

Four
(Improved)

Intersection Angle

Young’s
ModulusType of
soil

P

P

P

P

(Mpa)

Three models with intersection angle
0o , 60o, 90o
P

184

P

P

P

P

same module of elasticity in loading and unloading for

Surface Settlement(mm)Longitudinal Path of Shallow Tunnel(A-A)(m)
0
50
100
0

mohr-colomb model, when a shallow tunnel excavated heave
occurred. But in this article we study the effect of parameters
in settlement ratio and there is no significant error in our

5

ShallowBoth10

analysis.

15
20
25

Fig3.1. surface settlement of shallow tunnel alone andcrossing tunnels in path A-A
Mid Transverse Path of Shalloow
Tunnels(m)(B-B)
050
100
Surface Settlement(mm)

0
5
10

Deep
Fig 3.3. Bending moment in shallow tunnel crest during

Boththe excavation process

15

As we observe from Fig 3.3. before excavation and lining

20

installation, the bending moment is zero in intersection points
but after activation of lining, suddenly this moment reaches to25
Fig3.2. surface settlement of deep tunnel alone andcrossing tunnels in Path B-B
The first matter we get from longitudinal settlement ofshallow and deep tunnels(alone) is that the variation of
longitudinal settlement in distance 45 to 55 m (4.5D – 5.5D)very close to each other which describe the plane strain
conditions in model and this result could be used for
determining the minimum length of base model. However
this above condition is not valid in crossing tunnels and thenecessity of three dimensional simulations is clear. Withconsideration to figures 3.1 and 3.2, we understand that use of
linear superposition for estimating the total settlement of twotunnels construction is not true. In other word to gain the
maximum settlement, algebric sum of maximum settlementof shallow and deep tunnel construction, cause incorrect
outcomes.
A notable point from figure 3.2 that settlement of both tunnel
smaller than the settlement of deep tunnel only! Because of-30KN.m/m ( L point ) and in following process, this quantity
reduces because of lining installation in following process toreach a constant value, after starting the deep tunnel
excavation, increasing of moment continues until the deeptunnel approaches to intersection point then the bending
moment gradually increases and reaches to maximum value-42.6KN.m/m when deep tunnel crosses beneath the shallow
tunnel (M point). In other words when deep tunnel excavate
the increase of 36% in crest of bending moment in shallow
tunnel is observed. After crossing the deep tunnel beneath theshallow ones, the negative bending moment gradually
reduces until to reach the –11KN.m/m at the end of
excavation. The interesting object that bending moment
reaches a maximum value during the construction and thenecessity of stage construction analysis in soil structureinteraction observed.
With consideration to fig 3.1 and 3.2,we observed that thesettlement in initial and final 6 m of shallow or deep tunnel

185alone is equal to settlement of two tunnels. Also in diagram

excavation the quantity of settlement is equal to 26.6 mm that

3.3 it is observed that immediately after excavation of deepshows the 20% increase in maximum settlement. But when attunnel (stage 28) the moment doesn’t change in shallow

first the deep tunnel and then the shallow ones is excavated,tunnel crest but approximately after crossing the 6 m of deep

the produced maximum settlement is 30 mm in relation totunnel excavation, it increases gradually in shallow tunnel

previous condition we have 12% increasing in settlement.

crest . From these result we can conclude that if we far from

This approves that when settlement is at first importance, theintersection point about 4.5D the influence of deep tunnel on

shallow tunnel should be excavated first and then the deepshallow one can be ignore.

ones to produce the less Total settlement. In figure 3.5

This conclusion is important when you want to determine thebending moment in shallow tunnel crest has been shown by

area of influence.

three construction procedures.

In the future we study the influence of parameter oninteraction between tunnels.
3.1. Construction procedure
The first model is the base model which at first the shallow
tunnel and then the deep tunnel had been excavated. In second
model, two tunnels were excavated simultaneously thus,
when 3 m of shallow tunnel was excavated, in the following
process instead of next 3 m excavation, 3 m of deep tunnel
excavated and in the third model the deep tunnel excavatedinitially and then shallow tunnel excavated. For comparing
these three models, surface settlement in shallow tunnel width(path B-B in figure 2-1) have been shown in figure 3.4.

Figure 3.5. Bending moment of shallow tunnel crest in
Mid Transverse Path of Shallow Tunnel B-B(m)050Surface Settlement(mm)05three construction procedures

100

With consideration to figure 3.5 the minimum moment occurs

Shallow Then
Deep Excavatedin simultaneous excavation of two tunnels, but the directionof this moment changes alternatively (because of alternative

10
15
20
25
30

35

excavation of shallow and deep tunnel) afterwards the deep

Deep Then
Shallow
Excavatedtunnel excavation and then shallow ones give less bending
moment than shallow tunnel excavation then deep ones. Thus
the parameter of construction procedure acts differently inBoth Tunnels
Excavate
Simultaneously

settlement and moment. For example initial excavation ofshallow tunnel results in producing the less surface settlement

Fig3.4. influence of tunnels construction procedure on

but creates more bending moment in shallow tunnel crest.
The incremental ratio of bending moment (surface settlement of shallow tunnel

𝑀𝑀−𝑀𝐿
𝑀𝐿

) in

condition of shallow tunnel excavation and then deep ones
We get from figure 3.4 when at first the shallow tunnel and

63% increase, and in deep tunnel excavation and then shallow

then deep ones is excavated, the maximum of surface

ones 400% increase which shows the immense influence ofsettlement is 22.1 mm and during the tunnels simultaneous

construction procedure on bending moment alterations.

186

3.2. Soil characteristic

Transverse Path of Shallow Tunnel B-B(m)In this part according to table II, the influence of soil type on

0surface settlement and bending moment of shallow tunnel
Surface Settlement(mm)crest was investigated. So four models were produced withvarious soil type. In case four, soil is improved with groutingthat increase cohesion.With consideration to figure 3.6 and
3.7, the maximum settlement in first type soil during theshallow tunnel excavation is equal to 3.8 mm and when the
deep tunnel is excavated reaches to 10.5 mm, that indicate
maximum relative surface settlement 175% increase. Also
the maximum settlement in second type soil during theshallow tunnel excavation is 5 mm which this settlement after
deep tunnel excavation reaches to 14.1 mm that shows 180%
ralative surface settlement increase. In third type during theshallow tunnel excavation, the maximum settlement reaches
to 8 mm that after deep tunnel excavation underneath theshallow tunnel this quantity raise to 22.1 we understand the
175% relative surface settlement increase in intersection
point. In fourth type soil which is improved, settlement inshallow tunnel excavation is 1.6 mm which during the deeptunnel excavation reaches to 4.9 mm and shows a 212%
increase of maximum settlement during the shallow tunnel
excavation. So we cannot anticipate a distinct rule to compare
the interaction effect between soil type.

Surface Settlement(mm)

-2
0
2

100
Soil Type III

5

Soil Type II

10

Soil Type I
Soil Type IV

15
20
25

Fig 3.7.Surface linear settlement during the deep tunnel
excavation in different types of soils
According to figures 3.6and3.7 in shallow tunnel excavationfrom first type soil to second type and from first type soil tothird type results in 32 and 111 percent increase of settlement
respectively. In intersection tunnels from first type to second
type and from first type to third type soil results in 34 and 110
percent increasing of settlement. In other words the Soil
improvement in intersections area can result in settlement
reducing. When the soil is improved (fourth type soil) in
relation to first type in shallow tunnel excavation, themaximum settlement decreases to 58% and during the deeptunnel excavation the maximum settlement decreases to 52%.
In figure 3-6, the alterations of bending moment in shallow
tunnel crest during the excavation of this tunnel and deepTransverse Section of Shallow Tunnel B-B(m)
0
50
100
-4

50

0

ones have been shown.

With consideration to figure 3-6, in first type soil the moment
Soil Type III changes from -17KN.m/m(L point) to -28.KN.m/m(M point)
Soil Type II which shows 68% increasing in maximum bending moment.
Soil Type I

4For second type soil, the bending moment changes from
Soil Type IV -20KN.m/m to -35KN.m/m which shows 75% increasing in

6

bending moment during the excavation. Also for third type8

soil, the moment changes from-26KN.m/m to 42.4KN.m/m

10which shows 63% increasing in scale of bending moment. In

Fig 3.6. Surface linear settlement during the shallow
tunnel excavation in different types of soils

improvedsoil(fourth

type)

the

momentreaches

from-5KN.m/m to -10KN.m/m which shows 50% increasing.
A noticeable point is that in the case of the soil improvement
the relative bending moment decrease to a minimum value
compare to other soil type. The other important point we187realize here is that the soil improvement in intersection area

dimensional simulation can be used instead of three

not only causes the settlement decrease but also is a reason for

dimensional. In figure 3-8, the bending moment has been

decreasing the bending moment in lining tunnel, in contrast

shown in shallow tunnel crest. The noticeable point is thatwith other parameters that each of them if decrease theproduced bending moment in shallow tunnel crest in 90°

surface settlement increase the bending moment and viceintersection angle has the most quantity and in intersection

versa. But soil improvement decrease both of the surface

angle of 60°, this is approximately identical. So, a parameter

settlement and bending moment because in this case thenot only causes decrease in settlement but also increase inimproved soil carry more load than other soil type.

bending moment.

Surface Settlement(mm)Longitudinal Path of Shallow Tunnel(A-A) (m)
0
50
100
0
5
angle=9010angle=60

15angle=020
25
30

Fig 3.9. Surface settlement for three different intersection

Fig 3.8. Produced bending moment in shallow tunnel

anglescrest in different types of soils
3.3. Intersection angle
In order to clarify the influence of intersection angle between
two tunnels in surface settlement and bending moment, threemodels with angles of 60°, 90° , 0° have been made. The
surface settlement in length of shallow tunnel for three 60°,90°, 0° has been shown in figure 3-7. With consideration to
figure 3-7, the minimum settlement has been presented forintersection angle of 90°, the produced maximum settlement
for intersection angle of 0° hasn’t much difference withother settlement curves but this maximum settlement occurs
in all lengths of tunnel. As it described before the reason ofdifference in settlement in 24 m initial and final of tunnel for
0° is, this length was excavated with 3m Unsupported lengthand others with 1.5 m Also in angle of 0° we could observe

Fig 3.10. Bending moment in shallow tunnel crest forthree intersection angles

the plane strain conditions for parallel tunnels and the two188

4-Conclusion
Totally we can say if we go 4.5D farther from two tunnels
intersection point, the influences of interaction between twotunnels can be negligible.
Also with consideration to bending moment in presented
tunnel crest, the most bending moment occurs during theconstruction and when deep tunnel crossing beneath theshallow ones.
If a parameters cause to decrease the settlement, results inbending moment growth or vice versa but in soil
improvement, the improvement of soil transfers load to soil
and cause decrease in both settlement and moment. In general
the parameters change is more efficient in decreasing andincreasing of bending moment in proportion to decreasing or
increasing of surface settlement. Some parameters such as
unsupported length despite of increasing the relative
settlement decrease the total settlement.
5 - REFERENCES[1] Asano T, Ishihara M, Kiyota Y, Kurosawa H, Ebisu S.
An observational excavation control method for adjacent
mountain tunnels. Tunnel UndergrSpace Technology
2003; 18(2–3):291–301[2] Kim SH., 1996. Model testing and analysis ofinteractions between tunnels in clay. PhD thesis,
Department of Engineering Science, University of
Oxford.
[3] Liu HY, Small JC, Carter JP. Full 3D modeling foreffects of tunneling on existing support systems in the
Sydney region. Tunnel Underground Space Technology
2008; 23:399–420[4] Liu HY, Small JC, Carter JP, Williams. Effects oftunneling on existing support systems of perpendicularly
crossing tunnels. Computers and Geotechnics 2009;36:880-894.

189

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Use of post-flotation copper tailings in the construction of dump dams
Tschuschke W., Wierzbicki J.
Department of Geotechnics, Poznań University of Life Sciences, Poland
ABSTRACT
Considerable amounts of mine tailings are produced by the mining industry every day. Post-flotation tailings in themining of copper ores in Poland constitute 95% excavated mine run and are deposited in dumps. Constructional
elements of a dump include outer dams surrounding the object. Durability and stability of these building structures
determines safe operation of the entire object. Selection of adequate quality soils for the construction of dams is a
crucial factor limiting the risk of a potential disaster. Very large dumps require adequately large amounts of soil for theconstruction of dams. An alternative for this classical approach may be provided by the use of deposited tailings as the
structural material. In this solution the amount of mine waste deposited in the dump is reduced, there is no need to
exploit deposits of natural soils and the potential capacity of the dump is increased.
The paper presents principles of the construction technology of hydraulic fill dams, criteria for grain size distribution
and physical properties of sediments to be used in the construction of dams, principles of quality control for earth workand methodology of control tests. The paper comments on the effect of the applied technology on the condition of thenatural environment.
Keywords: mine tailings, dump construction

1. INTRODUCTION

2. THE OBJECT

Competent waste management in accordance with legal
regulations at the site of their deposition is a priority forproperly managed and exploited dumps. This problem
becomes particularly significant when deposited waste
products may be economically re-used. Such a situation
occurs e.g. in case of wastes classified as neutral to theenvironment. The category of such wastes frequently includes
also mining wastes, distinguished by two indicators: varied
composition, dependent on the site and technology of
produced tailings, and their considerable volume.
In open pit mines it is most typically overburden removed
from the layers over the deposit, while in deep mines it islandings coming from the construction of the mine andexploitation of the subsurface deposit. Thus formally thesuitability of the run of mine as engineering construction
material should be similar to that of natural soils. However, inpractice the input material comprises the run of mine after
processing, differing in terms of its strength from the material
deposited on the dump. A key element of the change resultsfrom the fact that the rock material of continuous structurechanges after crumbling and is transformed into groundmaterial with a three-phase structure. A very good example ofsuch material is provided by post-flotation tailings fromcopper ore treatment. The utilization of post-flotation tailings
as a construction material suitable for earthen structures
requires first of all the determination of physico-mechanicalproperties required in case of a construction material, whilesecondly we need to identify the use, type and parameters of
the structure, where the material is to be used.

Poland is a world leading producer of copper and silver,
exploiting non-ferrous metal ores located on thesouth-western part of the country. In that region the entire
copper ore mining and processing industry is found. The
winnings from all (solely deep) mines are subjected in theprocess line to mechanical treatment and flotation processes
and - after the isolation of copper headings - it constitutes
post-flotation tailings, which need to be properly managed. In
the global scale post-flotation tailings constitute approx. 95%
excavated deposit, which due to the volume of production,
results in the need to manage approx. 25 million Mg waste
annually. Starting from the year 1997 all post-flotation
tailings are transported to one dump, which recently has been
officially named the Żelazny Most mining refuse
neutralisation facility (OWOU). The amount of waste
deposited over the years of dump operation has resulted in theconstruction of a hydroengineering mega-structure, second(after the Chinese Great Wall) engineering structure seenfrom space.
The size of the object is indicated by its parameters: theamount of deposited tailings – over 500 million m3, dump
area – 14 km2, the capacity of the tailings pond - 7 million m3,
the length of dams surrounding the dump – 14.3 km, theheight of the highest dams – over 60 m (Fig. 1). Post-flotation
tailings are transported to the dump by hydraulic method, firstunder the pressure and then by gravity, in the form of a
mixture with process water from the mines, at mixture densityof 1.11÷1.15 Mg/m3. Refuse is discharged to the dump frompipelines located in the crown of external dams surrounding
the dump (Fig. 2). After discharge tailings particles are
sedimented on the dump beach, while water flows to the pond

190

located in the central part of the dump, where after clearing inflow towers it is directed to the secondary technologicalcycle. Guidelines for the safe operation of the dump require
the maintenance of the water line at a minimum distance of
200 m from the dam crown, which determines beach slope,naturally oscillating around 1%.

Fig. 1. The general view of the OUOW Żelazny Most
[KGHM Polska Miedź S.A.].
The dump is expanded by the upstream construction of thedam superstructure on hydraulically filled beaches. The
superstructure above the crown of star dams is constructedfrom tailings deposited on the beach [1]. In all cases theheight of the superstructure is 2.5 m and it is adapted to therotation system of feeder pipelines. Annual increment indump height is on average 1.3 m.

shape in deposits with a weaker binder. The basic componentof the fraction of less than 0.06 mm comprises chips ofcarbonate rocks, while in the coarser fraction quartz is thedominant material, found in considerable amounts in thefraction of 0.25÷0.1 mm. Shale content decreases with areduction of grain diameter. Characteristics of tailings
connected with their original geomorphology determine
mechanical properties of the material to be used in theconstruction of earth dams.
Tailings gravitationally silted up on the beach are subjected to
the natural process of sedimentation segregation, as a result of
which coarser fractions are deposited in the vicinity oftailings discharge sites, while finer fractions flow with waterover the surface of the beach towards the pond. The finest
tailings (carbonate-clay tailings) are excluded from the ringsilting process, as they may not be used in dam
superstructures. These wastes are discharged directly insidethe dump to seal the pond bottom. Hydraulic filling by
tailings is a cyclical process, in which a fragment of the beachis elevated by several dozen centimeters at a time. Then thebeach drying stage occurs, lasting up to 2 months, followedby another hydraulic fill process. The technology of tailings
hydraulic transport and beach hydraulic filling, the process ofdrainage, drying and beach formation cause significant
effects, i.e. spatial heterogeneity of sediment consolidation
and a specific macrostructure, characterized by horizontal
laminations. The number and thickness of laminations withthe smallest grain size increases with the distance from thetailings discharge site, limiting the range of the sedimentation
zone for deposits suitable for dam superstructure. Sediments
deposited in that zone as a construction material have to meet
specific compactibility criteria, and as a consequence - also
bearing capacity, since additionally they constitute the subsoil
for dam superstructures constructed by the upstream method.
A diagram for the dump section with an arbitrary division into
zones is presented in Fig. 3.

Fig. 3. Diagrammatic cross-section of dump: 1 – coarse
tailings (fine sand), 2 – intermediate zone (sandy silt), 3 –slimes (predominate silt and clay fraction) [3].

Fig. 2. Observed plunge pool at discharge point at the Żelazny
Most dump [3].
3. CHARACTERISTICS OF DEPOSITED TAILINGS
Post-flotation tailings produced from ores geneticallyconnected with dolomites and copper shales are characterizedby considerable grain angularity, while that from sandstones
have a milder morphology or even a preserved original grain

4. CRITERIA FOR THE SELECTION OF MATERIALFOR DAM CONSTRUCTION
Due to the necessity to regularly increase the dump capacity,
the processes of engineering design, tailings deposition anddam superstructures constructed are realized in parallel. The
adopted method of upstream superstructure development
results in a situation when only the start dams made fromnatural soils are founded on the subsoil. Successive stages ofsuperstructure development, proceeding as the dump is being191filled, are realized from drained sediments deposited onbeaches. Design of the dump development requires
geotechnical identification of deposited sediments.
Geotechnical parameters of sediments change with the
distance from the discharge site, while they also depend on
the type of rock from which they originate, applied
processing, disposal technology, filtration and consolidation
processes. Only some of the deposited sediments may be used
as a construction material in the creation of dam
superstructure. Extremely fine silty deposits exhibit
properties of a plastic material with limited shear strength.
Sandy deposits have the characteristics of a non-cohesivematerial with a much greater rigidity and shear strength,
while indexes referring to compactibility are best suited todescribe sediment behavior. Assuming the strength criterionas the starting point for the determination of suitability ofdeposits for construction purposes, a series of tests was
performed in a direct shear apparatus at deposit compaction
corresponding to maximum dry density from the Proctor test
(Fig. 4). Samples of deposits were differentiated in terms of
their grain size distribution, as identified by SFR (ratio ofsand to fines <0,074mm) [2]. The correlation identified on thebasis of model tests between internal friction angle and agrain size parameter SFR was used in the determination of thephysical criterion of sediment suitability (Fig. 5).

Fig. 5. Correlation between internal friction angle (φ) andSFR coefficient for compacted tailings within the range ofnatural moisture content equal ± 5% of optimum moisture
content.

Fig. 6. Grading envelopes for the Żelazny Most mine tailings.

Fig. 4. Standard Proctor test results for the Żelazny Most
mine tailings (ρ ds – maximum dry density, w opt – optimum
moisture content).
Relatively high values of the internal friction angles result
from angularity of sediment grains. On the basis of ananalysis of stability it may be assumed that sediments
embedded in dams, in order to meet the strength criterion(φ > 35o) at the required compaction, should have SFR > 2,
which corresponds to a limitation of the silt content in thesediments to 30%. The area of grain size variability insediments classified as suitable for dam construction ispresented in Fig. 6.

The identification of the distribution and potential predictionof changes in sediment grain size in the beach profile andcross-section constitutes the basic information in thetechnology of sediment collection from the beach for theconstruction of dams and for the determination of
geotechnical parameters of sediments for the needs of theanalysis of object stability. The process of sedimentdeposition results in a situation when the hydraulically filled
beach is a heterogeneous medium of random character.
Identification of deposit grain size, being a function of thelocation in space, may be considered as a deterministic
problem for a heterogeneous medium [3]. Probability ofoccurrence of sediments with varying grain size at a specificdistance from the dam may be investigated by clustering
sediments into groups with similar grain size or by theso-called continuous distribution method for one of the
adopted measures of the grain size curve (Fig. 7) [3], [4], [5].
It results from the analysis of Fig. 7 that the previously
determined criterion of granularity (fp > 70%) is met by
sediments deposited in the belt of the beach adjacent to thedam at a distance up to 70 m from the discharge site [3]. This

192

fragment of the beach was assumed as the potential zone ofearth work, from which collected sediments may be used fordam superstructures. An advantage of the earth work in theidentified zone of the beach is connected with the mixing of
sediments with different grain sizes, together with theelimination of laminations of cohesive deposits formed in thecourse of hydraulic filling and compaction of the subsoil forthe future dam superstructures.

Fig. 7. The probability of tailings grain size distribution withthe distance from the dam [3].
The conditions specified above for the construction of dam
superstructures from the material deposited in the dump, i.emine tailings, constituted the basis for the development of aprocedure defining the manner and sequence of operations:• Hydraulic filling of a specified fragment of the beachwith post-flotation tailings,• Drainage and drying of deposited sediments in thebeach to a moisture content corresponding to
optimal moisture content,• Using bulldozers, collection of hydraulicallydeposited sediments from a beach section of 70 m inwidth (at this stage the beach was further compacted
for the successive stages of dam superstructure
construction) (Fig. 8),• Compaction of the formed layer using a vibrating roll
until the required compaction criterion is met,
• Quality control testing of performed earth work.
A repetition of the entire cycle until the complete module ofdam superstructure, 2.5 m in thickness, is obtained, together
with the final geotechnical acceptance test.
5. QUALITY CONTROL OF PERFORMED EARTH
WORKS
The structures of high, hydraulically filled dams frompost-flotation tailings surrounding the object, which role is toaccumulate these wastes, requires strict adherence to thetechnological regime and particularly thorough quality
control tests of the performed earth work.
Meeting the criteria of grain size distribution and compactionof the material incorporated in the dam is a necessary
pre-condition for the safe operation of the object,guaranteeing its stability and durability. Within theframework of control tests, testing is performed on an
on-going basis by the surface method after the formation andcompaction of a successive layer of 0.5 m in thickness and thefinal verification tests by the depth penetration test after thecompletion of the construction stage of the dam
superstructure with a thickness of 2.5 m. Surface tests are
based on standard methods, i.e. isotope and volumetric tests(Fig. 9) [6].

Fig. 9. Troxler gauge and hand CPT equipment during testingtailings [6].

Fig. 8. The view of the zone of tailings used for dam
construction [3].
• Formation of a dam superstructure with a layer of up
to 0.5 m in thickness,

The former determines moisture content and density ofdeposits in the controlled layer and as a consequence - its
relative compaction – R (dry density – ρ d to maximum dry
density – ρ ds ) (Fig. 4). In the latter laboratory analyses ofphysical parameters are performed on the collected sediment
samples, which results - in relation to the Proctor test - makeit possible to determine the relative compaction andadditionally verify the grain size criterion. An original
solution in this respect is provided by the depth testing
method based on the standard cone penetration test CPT [3],[7]. In this method penetration characteristics are recorded,
i.e. cone resistance - q c and friction ratio - R f , which are193

indicators for the assessment of relative density – D r andgrain size in sediments over the entire analyzed profile.
Sediments embedded in the dams meet the compaction
criterion when the condition described by equation 1 is met.
for D r > 70%:
q c > exp((36.81∙ln(σ v0 ) + 733)/( 17∙σ v0 0.0876))(1)

where: σ v0 – overburden stress.
To verify measures of sediment compaction it is possible toevaluate the relative compaction - R by relative density – D r
on the basis of a correlation dependence expressed byequation 2.
R = 0,855 + 0.165·D r(2)

The required criterion for sediment grain size (f π < 30%)
expressed by CPT parameters is determined by condition R f <1.3% [3].
An example of an interpretation of CPT testing in order to
identify grain size and compaction of sediments in the dam
superstructure profile is presented in Fig. 10.

Fig. 10. CPT results at the background of quality criterion.
6. THE EFFECT OF DAM SUPERSTRUCTURETECHNOLOGY ON THE CONDITION OF THE
NATURAL ENVIRONMENTIn order to assure environmentally safe operation of theŻelazny Most mining refuse neutralization facility we need tomention the following crucial problems [1]:• Assurance of object stability,• Limitation of migration of contaminated processwaters from the dump to groundwaters,• Limitation of dusting.
The applied technology of gradual stage dam superstructure
construction from post-flotation tailings previously depositedin the dump, in case of the first two problems does not
generally differ from the conventional method of dam
construction. In case of a risk of stability loss in thethreatened sections of the dam the vent embankment of thedam is loaded. Limitation of process water migration from thedump to the subsoil is provided by the construction of ringdrainage at successive layers of the dump, girdling ditches
around the dump as well as the barrier of deep wells at the
dump forefield. In contrast to the previous aspects, the hazard
connected with dusting is a specific characteristic of theŻelazny Most facility. Post-flotation tailings deposited on the
dump beach and used in dam construction are characterized
by the capacity to dry rapidly. Dry sediments, free of thebinding force, are exposed to the action of wind and are easilytransported outside the dump area, forming fine dust
suspended in the air. The impact of dust on the environment is
so intense that as a counter-measure the surface of the beachand dams is covered with air-sprayed bituminous emulsion.
7. CONCLUDING REMARKSThe presented technology of gradual stage construction ofdam superstructure, performed in the course of the operationof the wet waste disposal site from selected material, i.e.
post-flotation tailings deposited in that dump, has at least one
spectacular economic advantage. The use of wastes for theformation of dams on the one hand reduces the amount ofwastes deposited in the dump, while on the other hand it
eliminates the need to exploit deposits of natural soils andtransport them to the crown of the constructed dam
superstructure. In case of a huge object such as the Żelazny
Most facility, in which the annual dam superstructure requires
approx. 300 thousand m3 soil, savings connected with theapplied technology covering two basic cost components, i.e.
elimination of fees for the disposal of sediments embedded in
the dams and no need to extract and transport natural soils fordam construction, reach over 2 million US dollars annually.
We also need to mention another advantage, i.e. the fact thatas a result of crushing of run of mine the resulting
post-flotation tailings have the characteristics of fine-grained
broken aggregate, easily compacted and with advantageous
shear strength parameters, provided that the strict
technological regime of the dam superstructure construction
is maintained. A problem which may be perceived as a
drawback of this technology is the adverse environmental
impact connected with increased dusting of dried sediments
embedded in the dams. This may be solved by the applicationof appropriate counter-measures reducing dust emissions toareas adjacent to the dump.
8. REFERENCES[1] Wolski W. “Tailings dams – selected question”, Proc. of
13th European Conference on Soil Mechanics andGeotechnical Engineering, Prague, Vol. 3, Vanicek et al.
(eds.), pp. 167-176.
[2] Lipiński M.J., Wolski W., Fioravante V., Jamiolkowski
M.B., “Preliminary evaluation of hazard due toliquefaction for Żelazny Most tailings pond”, Proc. of
14th International Conference on Soil Mechanics andFoundation Engineering, Vol. 3, Hamburg 1997, pp.
1843-1846.
[3] Tschuschke W., “Cone penetration tests in-post
floatation sediments”,
Silesian University of194

Technology, Civil Engineering (110), Gliwice, Poland
2006, p. 266.
[4] Pordzik P., Tschuschke W., Wierzbicki J., “Statisticalevaluation of tailings grain size distribution of theŻelazny Most Reservoir dams”, Proc. of 12th Polish
National Conference on Soil Mechanics and Foundation
Engineering, Szczecin-Międzyzdroje, Poland 2001,Vol.
1b, pp. 181-191.
[5] Gołębiewska A., Lipiński M.J., “Measures forcharacterization of grain size distribution of post
floatation tailings”, Proc. of 12th Polish National
Conference on Soil Mechanics and Foundation
Engineering, Szczecin-Międzyzdroje, Poland 2001,Vol.
1a, pp. 211-219.
[6] Wierzbicki J., Niedzielski A., Waliński M., Wołyński
W., “The quality control of postflotation reservoir dam
by determination of relative compaction index in various
methods”, Geotechnical and Geophisical Site
Characterization Viiana da Fonseca & Mayne (eds.).
Millpress, Rotterdam 2004, ISBN 90 5966 009 9, pp.
1371-1375.
[7] Młynarek W., Tschuschke W., Welling E., “Control ofstrength parameters of tailings used for construction ofreservoir dams”, Proc. of the 5th International Conferenceon Tailings and Mine Waste, Fort Collins, Colorado,
USA 1998, pp. 213-221.

195

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effects of Slope Inclination on the Rain-induced Instability of Embankment
Slopes
Chaminda Gallage1, Shiran Jayakody1 and Taro Uchimura2
1

Queensland University of Technology, Brisbane, Australia
2
The University of Tokyo, Tokyo, Japan .

ABSTRACT
Rainfall has been identified as one of the main causes for embankment failures in areas where high annual rainfall is
experienced. The inclination of the embankment slope is important for its stability during rainfall. In this study,
instrumented model embankments were subjected to artificial rainfalls to investigate the effects of the slope inclinationon their stability. The results of the study suggested that when the slope inclination is greater than the friction angle ofthe soil, the failure is initiated by the loss of soil suction and when it is smaller than the friction angle of the soil, thefailure is initiated by the positive pore water pressure developed at the toe of the slope. Further, slopes become more
susceptible to sudden collapse during rainfall as the slope angle increases.
.
Keywords: Embankment model tests, instrumentation, soil suction, transient seepage, rain-induced slope failures

1. INTRODUCTION
Embankments are very useful geo-engineering structures thatare widely used to support roads and rails. They are
constructed by shaping the natural slope or by compacting theimported soils. Failure of these embankment slopes can cause
for human casualties as well as negative impacts on acountry’s economy. Most embankment slopes are initiallyunsaturated prior to any rainfall. In addition to triggering by
earthquakes, rainfall has caused the failure of many
embankment slopes, worldwide [1]-[4].
Reference [5] performed a series of parametric studies tounderstand the significance of hydrological and geotechnicalparameters of a slope on its rain-induced instability. The
study revealed that rainfall, rainfall intensity, soil properties,
location of the ground water table, and the slope geometry(angle, height) play a significant role in the rain-induced
instability of a slope. Angle and height of an embankment are
two key geometrical parameters that control the extent of thearea covered by the embankment slope and its stability. In
developed/developing area areas, there is a great demand to
reduce the area covered by the embankment slopes and hence
the embankments are constructed with steeper slopes and insome occasions geo-reinforced embankments are constructedwith vertical slopes [6].
Unsaturated embankment slopes can be stable with very steep
slope angle until it is exposed to rainfall. The mechanism ofslope failure induced by rainfall can vary depending on theslope angle. It is important to understand the mechanism andinitiation of failure of an embankment slope to provide bettercountermeasures against the failure. This paper presents theresults of the embankment model tests conducted toinvestigate the effects of slope angle on the mechanism andthe initiation of the failure the of embankment slope subjected

to rainfall. Two instrumented model embankment slopes withtwo different slope angles (e.g. 300 and 600) were subjected toartificial rainfall until the failure.
2. TESTING MATERIALS
Edosaki soil procured from a natural slope in Ibaraki
prefecture in Japan was employed in the experimental workof this study. Wet sieving and hydrometer analyses were
performed on this material as it contains fines (particles finer
than 0.075 mm) content of 17.1 %. The grain-size distribution
curve of the test material is shown in Figure 1. The specificgravity, maximum void ratio, and minimum void ratio of thesoil were measured as 2.75, 1.59, and 1.01, respectively. The
soil was found to be non-plastic. The optimum gravimetricwater content and the maximum dry density of the sandobtained from the standard proctor compaction test were
16.5% and 1.725 g/cm3 respectively. According to the
Unified Soil Classification System, the soil can be classified
as silty sand2.1 Soil Water Characteristic Curves (SWCCs)
Figure 2 depicts the soil-water characteristic curve (SWCC)
for Edosaki sand obtained in the laboratory using a Tempepressure cell. Both the drying and the wetting SWCCs wereobtained for the test material using a sample with dry densityof 1.22 g/cm3. Model slopes were constructed using Edosaki
sand to achieve dry density of 1.22 g/cm3.Since the air-entry
value of the ceramic disk used in Tempe pressure cell is 300
kPa, the measured SWCCs were restricted to the maximum
suction of 200 kPa. More details about SWCC measurement
using Tempe pressure cell is given in [7]. As shown Figure 2,
the laboratory measured SWCC data were fitted using theequation proposed by Fredlund and Xing [8].

196

(c)

Coefficient of permeability,k, [cm/sec]

Percent finer [%]

100
80
60
40
20

0.005

0.003
0.002
0.001
0.000
0

0
0.001

Dry density = 1.22 g/cm3
Experimental (Drying)
Prediction-Drying ( Fredlund & Xing)
Prediction-Drying ( Green & Corey)
Prediction-Drying ( Van Genuchten)

0.004

2

4

6

8

10

Suction, ua-uw

0.01

0.1
1
Grain size [mm]Figure 3. Measured and predicted permeability functionof test material

10

Volumetric water content

Figure 1. Grain size distribution curve of Edosaki soil
Dry density= 1.22 g/cm3
Experimental ( Drying)Experimental ( Wetting)
Fitting by Fredlund & Xing ( Drying & wetting)
Fitting by Fredlund & Xing ( Average))

0.4
0.3
0.2
0.1
0.0
0.1

1

10
Suction [kPa]100

Figure 2. SWCCs of test material
2.2 Unsaturated Permeability
Permeability function of unsaturated soils was measured
using a laboratory developed permeameter based on steady–state method [9]. Figure 3 shows the variation of measured
coefficient of permeability with suction following the drying
path of the SWCC. The test was conducted on the dry densityof 1.22 g/cm3 of Edosaki sand. Figure 3 also compares thelaboratory measured permeability function with thepredictions using the methods proposed by Fredlund [10],Green & Corey [11], and Van Genuchten [12] (all threemethods are included in SEEP/W -2004).
2.3 Unsaturated Shear Strength
A conventional direct shear apparatus was modified tomeasure unsaturated shear strength under controlled suction.
Figure 4 (a) depicts the variation of the φ’ with the suction.
The φ’ was obtained from both saturated and unsaturated
consolidated drained tests employing the modified direct
shear device. The results suggested that the effects of suction
on the φ’ is not significant. However, as shown in Figure 4(b), the apparent cohesion (c) increases with the suction in adecreasing rate.

Figure 4. Measured unsaturated shear strength properties oftest material
3. MODEL TESTSIn order to examine the effects of slope inclination on themechanism of rain-induced slope (embankment) failures, two
instrumented model tests with two different slope inclinations(i.e. 300 and 600) were subjected to an artificial rainfall until
the failure of the slope. Pore-water pressures, volumetricwater contents, soil displacements were monitoredcontinuously at different locations on and in the slope during
the construction of the slope and the application of therainfall.
3.1 Instruments Used
The tank (Figure 5) used in model test has a length of 2.0 m,
width of 0.8 m, and a height of 1.0 m. The walls of the tankare made up of steel plates, except for the front side which ismade of acrylic glass for easy observation of deformation
processes.
ADR (Amplitude Domain Reflectometry) and ECHO types
soil moisture sensors were used in the model tests to measure197

soil moisture content during water infiltration. In order tomeasure both positive and negative water pressures,
KYOWA 05 PMG pressure sensors were modified with aceramic cup of 100 kPa air-entry value. Figure 6 shows thetypes of sensors used in the model tests. LVDTs were used tomeasure the local displacements on the slope surface.
Inclinometers installed in the soil measured the sub-soil
displacements. Evaflow side spray irrigation tube (hole size,
0.1 mm) was used to simulate rainfall as shown in Figure 5.

Figure 5. Tank used for model tests

during rainfall application.
3.3 Rainfall Simulation
Approximately 24 hours after the completion of theconstruction of the slope, the slope was subjected to a rainfallwith an approximate intensity of 40 mm/hr. During therainfall application, it was set to log data for every 2 seconds
from all the sensors connected with the logging
system. Evaflow side spray irrigation tube system used in thisexperiment was able to provide a relatively uniform constantrainfall between 20 mm/hr and 50 mm/hr. It was able tocontrol the rainfall intensity by regulating the pressure of thewater supplied to the system. This rainfall system wascalibrated to give different rainfall intensities with supplied
water pressure. Four measuring cylinders (rain gauges) wereused to measure intensity of rainfall; two of them werepositioned near the toe and the other two were placed on the
top of the embankment. The rainfall intensity mentioned inthis study was calculated by averaging the measured rainfall
over the entire duration of rainfall and the numbers ofmeasuring points. To minimize the possible fluctuation of
rainfall intensity with time and space (due to wind), therainfall was applied in the morning (e.g., 5.30 a.m) in mosttests.
4. RESULTS AND DISCUSSIONS

Figure 6. Sensors used in model tests
3.2 Model Preparation
Two model embankments with different slope inclinations(e.g. 300, 600) were constructed by compacting Edosaki sandwith the initial gravimetric water content of 13.5 %. The
compaction was performed in order to achieve the dry densityof 1.22 g/cm3. 15 pore-water pressure sensors, 15 watercontent sensors, three inclinometers, and two LVDTs were
installed in each model embankment. All the sensors wereconnected to a data acquisition system for continuous logging
of data during the construction of the slope and as well as

Figure 07 shows the locations of the different transducers(pore-water pressure, water content, and horizontal
displacement) installed in the model slope which has an
inclination of 30o. At the beginning of the artificial rainfall of
40 mm/hr, the slope was in unsaturated conditions with thesuction of 6 ~ 12 kPa (Figure 8(a)) and corresponding
volumetric water content 0.2 ~ 0.15 (Figure 8(c)). As therainfall progressed, the water contents and the pore-waterpressures at the different locations in the slope increased asshown in Figure 8(b) and (a), respectively. The lower part ofthe slope became saturated first forming a water table below
the toe area of the slope. As shown in Figure 8(c), thehorizontal deformation of the slope started when the toe of theslope was saturated (P8 registered positive value by the timethe deformation started) at about 5000 sec after the beginningof rainfall. The deformation seemed to be progressive (Figure9) with the rise of water table. The rate of horizontal slope
displacement was measured to be 3.4 ~ 7.9 mm/hr.
A slope with an inclination of 60o was constructed andinstrumented as shown in Figure 10. Once the slope wassubjected to an artificial rainfall of 45 mm/hr, as shown in
Figure 11(c), the horizontal deformation of the slope (D1, D2,
and surface at inclination 3 (Incli.3)) began with a rate ofabout 14.4 mm/hr even before the water table rose up to thelevel of P10 and P8 (Figures 11(a) & (b)). Cracks appeared at
the top (crest) of the slope with the start of deformation (atabout 2000 sec after the beginning of rainfall). Within 1000
sec after the cracks appearance at the top, the entire slope
collapsed as shown in Figure 12. It was interesting to observethat the most part of the slope was unsaturated (the water front

198

did not reach some areas) by the time the entire slope
collapse. It seemed that this failure was initiated by thedecrease in the shear strength of the slope due to the loss ofsoil suction. An unsaturated slope with the inclination greaterthan the friction angle of the soil can be stable until thesuction is decreased by the rain water infiltration into theslope.
Pore pressure sensor [P] with water content sensor [M]200(All dimensions in mm)

100

400

M1-M8 are ADR probs
M9-M15 are ECHO
P5M5

P1M1

100
50
50
100
P11M11

100

P14M14

100
50

er 3

Inclinomet

P4M4
P12M12

Figure 9. The progress of the failure of the slope with 300
inclination

Inclinometer 1&2

D1

P3M3

Perforated wall

P6M6
P2M2

P15M15

P7M7
P13M13

50
200

200

0

P9M9

100

300

250

350

P10M10

30

200

250

P8M8

D2
100

Figure 7. Sensor locations in model test with 300 inclination

Pore-waterpressure [kPa](a)Volumetric Water
Content,θ(b)Horizontal
Displacement [mm]

(c)

0
-4

P1
P2
P6
P14
P8

-8
-12
0.4
0.3

M1
M2
M6
M14
M8

0.2
0.1
8
D2 (Toe of the slope)
Incli.3 (Middle of the slope)
D1 (Near Top of the slope)

4
0
0

2000

4000

6000

Time [sec]Figure 8. Time histories of: (a) pore-water pressure; (b) volumetric water content; and (c)displacement at various locations of the slope with 300inclination
199

8000

Pore pressure sensor [P] with water content sensor [M]200(All dimensions in mm)
400

100

P5M5

M1-M8 are ADR probs
M9-M15 are ECHO

P1M1

100

P3M3

100

Inclinometer 1&2

D1

P12M12

100

P2M2

er 3

Inclinomet

P11M11

100

P6M6

100

P15M15

P4M4

Perforated wall

P9M9

D2

0

100
50
50

P14M14200

P13M13

P7M7 60

P8M8100 100 100 100

200

P10M10100

200

Figure 10. Sensor locations in model slope with 600
Inclination

Figure 12. The progress of the failure of the slope with 600
inclination

Pore-waterpressure [kPa]

(a)
0
-4

P1
P2
P6
P13
P8

-8
-12
-16

Volumetric Water
Content,θ(b)Horizontal
Displacement [mm]

(c)

0.4
M1
M2
M6
M13
M8

0.3
0.2
0.1
8
D2 (Toe of the slope)
Incli.3 (Middle of the slope)
D1 (Near Top of the slope)

4
0
0

1000

2000

3000

Time [sec]Figure 11. Time histories of: (a) pore-water pressure; (b) volumetric water content; and (c)displacement at various locations of the slope with 600inclination
200

4000

5. CONCLUSIONS
The following conclusions can be drawn from thisexperimental investigation:• Slopes become more vulnerable to sudden collapse
with the increase in their inclinations.
• Failures of gently slopes are initiated by thedevelopment of positive pore-water pressure near
the toe and the failure is relatively slow andprogressive.
• In the case of steep slopes (especially when slope
inclination is greater than the soil’s friction angle),
the decrease in suction is a predominant factor for
the failure initiation.

[12] Van Genuchten, M. T. (1980). “A closed-form equation
for predicting the hydraulic conductivity of unsaturated
soils.” Soil Science Society of American Journal, 44:892-898.

REFERENCES[1] Kunio, K. and Hiroakai, S. (1989). “Embankment designin mountainous area with heavy rainfall.” Soils andFoundations, 29(4), 35 – 48.
[2] Okada, K. and Sugiyama, T. (1994). "A risk estimation
method of railway embankment collapse due to heavy
rainfall." Structural Safety, 14(2), 131-150[3] Aleotti, P. (2004). “ A warning system for
rainfall-induced shallow failures.” Engineering Geology,
73(4), 247 – 257.
[4] Ng, C.W.W. and Zhan, L.T. (2007). “Comparative studyof rainfall infiltration into a bare and a grassed
unsaturated expansive soil slope.”
Soils andFoundations, 47(2), 207–217.
[5] Rahardjo, H., Ong, T. H., Rezaur, R. B., and Leong, E. C.
(2007). “Factors controlling instability of homogeneous
soil slopes under rainfall.” Journal of Geotechnical andGeoenvironmental Engineering, 133 (12), 1532–1543.
[6] Tatsuoka,
F.
(2011).
“Developmentsofgeosynthetic-reinforced soil technology in Japan:
Retaining walls and bridge abutments”. Proceeding ofFirst International Conference on Geotechnique,
Construction Materials, and Environment, Mie, Japan.
[7] Gallage, C. P. K., and Uchimura, T. (2010). “Effect ofdry density and grain size distribution of soil-water
characteristic curve of sandy soils.” Soils andFoundations, 50(1), 161-172.
[8] Fredlund, D. G., and Xing, A. (1994). “Equation for thesoil-water characteristic curve.” Canadian Geotechnical
Journal, 31, 521-532.
[9] Klute, A. (1965). “Laboratory measurement of hydraulicconductivity of Unsaturated soil.” In methods of soil
analysis, Mono. 9, Part 1, Amer. Soc. Of Agronomy,
Madison, WI, 253-261.
[10] Fredlund, D. G., Xing, A., and Huang, S. 1994.
Prediction of the permeability function for unsaturated
soils using the soil-water characteristic curve, Canadian
Geotechnical Journal, 31: 533-546.
[11] Green, R. E., and Corey, J. C. (1971). “Calculation of
hydraulic conductivity: A further evaluation of somepredictive methods.” Soil Science Society of America
Proceedings, 35, 3-8.
201

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Extension of an integrated AHP and TOPSIS approach in the earth dam site
selection
Minatour Yasser1, Khazaie Jahangir2,*
1. M.Sc. Student of Water Eng., Dept. of Civil Eng., Razi Univ., Kermanshah, Iran
2. Assist. Prof. of Civil Eng., Dept. of Civil Eng., Razi Univ., Kermanshah, IranJ.khazaie@razi.ac.ir
ABSTRACT
Optimal locating the dam site is one of problems associated with water resources management and it depends on aset of qualitative and quantitative criteria. Therefore, such problems can be resolved by multi-criteria decision making(MCDM) approach. This study aims to develop MCDM approach to locate the dam site, for this purpose an integrated
AHP (Analytic Hierarchy Process) and TOPSIS (Technique for Order Performance by Similarity to Ideal Solution)
approach was applied in the Harsin earth dam site selection at the western part of Iran. To do this, firstly the influential
criteria for locating the earth dam site were determined using comprehensive literature review and the experts’opinions. Then, some watersheds in the surrounding areas of Iran’s Harsin city were studied and four feasible sites
proposed. In the final stage, these sites in order to construct a multi-purpose earth dam were prioritized using thisintegrated approach, and the most optimal site was selected. AHP was used to determine the relative weight of criteria
and TOPSIS was used to rank the alternatives.
Keywords: AHP; Earth dam; Multi-Criteria Decision Making (MCDM); TOPSIS, Site selection.
1 INTRODUCTION
Water resources management because of the water crisisand depending on various factors is one of the mostimportant and complex problems in different countries. The
dam site selection is one of problems associated with water
resources management. In addition, dam construction inflicts
high costs and it should be constructed in a site with higher
potential to compensate the costs. In this situation in order tolocate the optimal dam site, various studies must be
conducted. As this issue is related to different qualitative andquantitative criteria, decision making in this area is complex
act. Therefore, in order to solve this problem, multi-criteria
decision making (MCDM) methods could be used. MCDM
is one of the branches of operation research whichinvestigates the decision making problems under somedecision making criteria. This kind of decision making
utilizes a number of criteria instead of solely one optimalassessment criterion. MCDM is categorized into two maingroups of multi-objective decision making (MODM) andmulti-attribute decision making (MADM) (Pohekar andRamachandran 2004). The aim of the MODM problems isthe simultaneous optimization of a number of targetfunctions under a series of limitations. Generally, MODM
techniques are applied to design and optimization problems.
The MADM methods are used for selecting the bestalternative amongst various alternatives and obtaining aranking for the alternatives. different methods have beenpresented for MADM, including: compromise programming[2]; elimination et choice translating reality I (ELECTRE I)[3,4], ELECTRE II [5], and ELECTRE III [6]; analytic
hierarchy process (AHP) [7]; simple additive weighting(SAW) [8]; technique for order preference by simulation ofideal solution (TOPSIS) [8]; preference ranking organization
method for enrichment evaluations (PROMETHEE) [9];gray relational analysis [10]; the weighted sum method(WSM) [11]; visekriterijumska optimizacija i kompromisno
resenje (VIKOR) [12,13]; and data development analysis[14].
MADM methods have been used in the various fields ofwater engineering [15-30]. But, the application of thesemethods to locating the earth dam site has not been reported
yet. In this study, an integrated AHP and TOPSIS techniquewere used to locate the earth dam site.
2. METHODOLOGY
In this study, the AHP introduced by saaty (1980) wasused to determine the weight of criteria and TOPSIS firstdeveloped by Hwang and Yoon (1981) was used to rank thealternatives. Decision making using this model involves two
essential stages: applying the AHP to determine the weightof criteria, applying the TOPSIS method to rank alternatives.
2.1. Applying the AHP to determine the weight of criteria
The AHP is one of the most powerful and simplest
techniques for solving MADM problems. The working
principle of this technique is based on the judgment of theuser and pair-wise comparisons. Weight of criteria using thismethod is calculated in two steps as follows:2022.1.1 Construction of pair-wise comparison matrix
In the first step, pair-wise comparison matrix of criteria
is constructed as:

w1
w
 1
w2
wn

w2
w2 

w2
wn 
   

wn
wn 

w2
wn 

 w1
w
 1
 w2

A =  w1


 wn
 w1

chosen alternative should have the shortest distance from the
Positive Ideal Solution (PIS) and the farthest from the
Negative Ideal Solution (NIS) for solving MCDM problems.
In short, the PIS is composed of all the best indices, whereas
the NIS is made up of all the worst attainable indices.
Assume a MCDM problem has m alternatives (A 1 , A 2 , . . . ,A m ), and n decision criteria (C 1 , C 2 , . . . , C n ). Eachalternative is assessed with respect to the n criteria. All thevalues allocated to the alternatives with respect to eachcriterion form a decision matrix shown by X = (x ij ) m×n . Let
W= (w1 , w2 , . . . , wn ) be the relative weight vector of the
R

R

(1)

R

R

R

R

R

R

R

R

R

R

R

criteria, satisfying

Where A is the comparison pair-wise matrix of criteria; w1
the weight of criterion 1; w2 the weight of criterion 2 and wn
is the weight of criterion n.
To determine the relative preference for two criteria of
the hierarchy in matrix A an underlying semantic scale isemployed having values from 1 to 9 according to Table 1.
Table 1 Saaty’s scale of preferences in the pair-wisecomparison process [7]R

Definition

1

Equal importance
Weak importance of one over
Another

5

Essential or strong importance7

Demonstrated importance9

Absolute importance

2, 4, 6, 8Intermediate values between
adjacent scale values

R

R

R

R

R

R

∑ w = 1 . The procedure of TOPSIS
n

j =1

j

can be expressed in a series of steps [31-34]:
2.2.1 Calculate the normalized decision matrix
The normalized value n ij is calculated as:
R

xij

nij =

∑i=1 xij2mR, i = 1,2,3, , m; j = 1,2,3, , n(6)

2.2.2. Calculate the weighted normalized decision matrix
The weighted normalized value v ij is calculated as:
R

Intensity of Importance
3

R

R

R

vij = w j nij , i = 1,2,3,  , m; j = 1,2,3, , n(7)

Where, wj is the weight of the jth criterion.
RR

2.2.3 Determine PIS and NIS
The PIS indicates the most preferable alternative, and the
NIS indicates the least preferable alternative. So determine
PIS and NIS as follows:{}

A+ = v1+ ,  , vn+ = (max vij i ∈ E ), (min vij i ∈ F )
j
 j

(8)

2.1.2 Calculation of relative weights
In AHP, the eigenvector method uses for calculating therelative weights. In this method, firstly, by using Eq. 2 theeigenvalues of matrix A are calculated. Then, the relative
weights of matrix A (the relative weights of criteria) are
obtained with simultaneous solution of Eqs. 3 and 5:(2)
AW = λW( A − λmax I )W = 0(3)

W = ( w1 , w2 , w3 , ..., w j , ..., wn )(4)

∑ w =1

(5)

n

j =1

j

Where λ max is the biggest eigenvalue of matrix A, W theweight vector, wj weight of criterion j and I is the unit
matrix.
R

P

P

P

P

2.2.4 Calculate the distance from PIS and NIS.
The Euclidean distance of each alternative from the PIS
is obtained as:nd i+ = ∑ (vij − v +j ) 2  , i = 1, 2, 3,  , m j =1R

2.2 Applying the TOPSIS method to rank alternatives
Among the MCDM methods, TOPSIS is a practical anduseful technique for ranking and selecting a number ofpossible alternatives via measuring Euclidean distances. The
working principle of TOPSIS is based on the fact that the}(9)
Where, A+ and A- are PIS and NIS, respectively. E isassociated with benefit criteria (+), and F is associated withcost criteria (–).

R

R

2T

{

A− = v1− ,  , vn− = (min vij i ∈ E ), (max vij i ∈ F )
j
 j


(10)

Similarly, the Euclidean distance of each alternative fromthe NIS is given as:

2T

203


n
d i− = ∑ (vij − v −j ) 2  , i = 1, 2, 3,  , m j =1(11)2.2.5 Calculate the relative closeness of each alternative
to the PIS
The relative closeness of the alternative A i with respect
to A+ is defined as CC i :

CCi =

d i−
, i = 1, 2, 3, , m
d i+ + d i−

(12)

2.2.6. Determine the overall ranking
The overall ranking of alternatives is obtained using therelative closeness to the PIS. Each alternative with higher
the CC i has the higher priority.
3 CRITERIA
In order to determine effective criteria in selecting anappropriate earth dam site, comprehensive literature review
were conducted and the most important criteria was selected.
A brief explanation about the criteria is presented:
3.1 Health dam site (C1)
The dam site must be located in a place where seams andtracks and the possibility of tectonic activities such as anearthquake, landslide, subsidence, etc., are minimal andabutments are more stable. Also, with lesser permeability andliquefaction properties of soil and natural materials in the damsite, better results will be achieved. In addition, soil
mechanical properties of the region (compaction,
consolidation, etc.) and kind of region geological layers interms of influence on reservoir water quality (for instance, it
must not be limestone), construction of dam reservoir andfoundation, stability in the condition of rapid discharge of thereservoir, etc., should be considered.
3.2 Overall cost (C2)
The overall cost for the construction of the dam is anotherimportant criterion in selecting the earth dam site. Thisoverall cost is includes: construction of the dam body andreservoir, water diversion during construction, water transfer
to the consumption location, energy supply, site preparation,
land use, and other costs associated with the dam projects.
3.3 Annual yield (C 3 )The annual volume of the water that passes through thecross section of the river in the dam site is called the annual
yield, which plays important roles in locating the dam site.
3.4 Topographic conditions (C4)The existence of the secondary valley or stone abutmentswith suitable topography around the main river forconstructing dam spillway is important. Also, when the mainriver is U or S shaped, it makes the length of tunnel, channel,
and other water transfer systems short to divert or transfer
water from upstream to downstream at the time of the dam
construction and after that. In general, with regard to dam
reservoir and its body, the best site for the dam is a place
where a wide valley with high walls leads to a narrow canyon
with tenacious walls.

3.5 Access to materials and facilities (C5)
Access to materials (borrow sources, cement, etc.) andfacilities (the power transmission lines, oil and gas
distribution pipelines, road, etc.) is important.
3.6 Economic Development (C6)The effect of dam construction on agriculture andindustrial development, power generation, fishery, job
creation, etc., which are related to economic development, areregarded as important attributes for selecting the dam site.
3.7 Water quality (C7)
The quality of water stored in reservoirs used for drinking
and agriculture is important. Therefore, the dam site must be
selected in a place where the quality of the water entering thereservoir is better. Also, factors that may reduce water quality(such as salt mine, gypsum, lime, etc.) should be considered.
3.8 Damage of dam body and reservoir (C8)The damages caused by the construction of the dam bodyand reservoir, including: environmental damages, destroying
mines, historical monuments, agricultural lands andresidential areas; displacement of roads, Railway and powerlines, change the route of oil and gas pipelines,
telecommunication facilities, etc. should be considered.
3.9 Volume of reservoir (C9)When the reservoir created after dam construction has
more volume, the surface area of the reservoir water isincreased, and it has more impact on the climate, but
evaporation and water pollution potential increase. On theother hand, if the dam is to be constructed in a site where byincreasing the volume of reservoir water the surface area ofthe reservoir water will not significantly be altered (the slopeof the valley is high), the height and water hydrostaticpressure will be increased, and it is beneficial in terms ofenergy generation and water transfer downstream. But, more
power would be imposed on the dam’s body, and the structure
should be built higher and stronger. Therefore, the dam
should be constructed where the reservoir capacity based on
the above-mentioned items, leakage, and other losses are
optimal.
3.10 River flow regime (C10)Permanent or seasonal flow regime of the river at the damsite is important. The seasonal rivers have more sedimenttransport and less water quality, and also, the management ofwater resources is harder due to the lack of accurate dischargeof water entering the reservoir. Therefore, a permanent flow
regime is recommended.
3.11 Water diversion and transfer (C11)
Dam site must be selected in a place where the costs ofwater diversion during construction an water transfer toconsumption location are lesser. Also the costs ofmaintenance of water diversion and transfer systems (such as:channel, tunnel, pipe, etc) and other costs and damages (such204as: environmental, displacement of roads, railway and powerlines, change at the route of oil pipelines, gas,
telecommunication facilities, etc) are lesser.
3.12 Annual volume of sediment (C12)
When the annual volume of sediment entering thereservoir is lesser, the volume of the reservoir during its
useful life, water quality, and the efficiency of the dam arehigher.
3.13 Probable dam break (C13)
Material and moral damages caused by a probable dam
break are among the important attributes for locating the damsite, and the dam should be constructed in a place where theamount of damages due to its probable break is low.
3.14 Probable maximum flood (C14)The maximum volume of water caused by thawing snow
and ice or other atmospheric precipitation occurring probably
in rivers is called probable maximum flood.
3.15 Average annual evaporation (C15)Regarding the annual average temperature difference in
Iran’s different regions, evaporation from the dam reservoir
in different areas is different. And this affects the retention
time of water in the reservoir (in terms of volume) andconsequently the efficiency of the dam.

Fig. 1 Location of Harsin city in Kermanshah province, IranThen, according to the effective criteria in the dam site
selection (mentioned in the previous part) and using theexperiences and opinions of experts and existing informationand data, four feasible alternatives in the surrounding areasof Harsin city were proposed, which are known as letters‘A’, ‘B’, ‘C’ and ‘D’. Fig. 2 shows the location of these
proposed alternatives in studied field.

3.16 Environmental impacts (C16)
Changing weather conditions, vegetation, and wildlife areother factors that play significant roles in locating the damsite.
3.17 Social impacts (C17)The The social impacts of population center displacement
and the integration of different ethnic cultures due to thedestruction of residential places for dam construction,
reservoir dewatering, and also the utilization of the dam water
downstream should be considered.
3.18 Political impacts (C18)
The dam construction goals for reducing political tensions
including water supply of a city, preventing grievances andimmigration of residents of a border city, etc., are among theattributes that should also be considered.
4 CASE STUDYThis study aims to develop an integrated AHP andTOPSIS approach to locate the dam site, in this section,
presented methodology was applied in the optimal selectionof the Harsin earth dam site. The purpose of this dam isdevelopment of agriculture and Industrial, drinking watersupply, power generation, fishery, etc. in Harsin city at thewestern part of Iran. This city is located 16 km east of
Kermanshah city (the capital of Kermanshah province, Iran).
Its location is 34°16′ longitude and 45°35′ latitude.
Fig. 1 shows the location of Harsin city in Kermanshah
province, Iran.

Fig. 2 Location of the proposed alternatives in studied field
After selecting criteria for locating the earth dam site andproposed alternatives (see Fig. 2), integrated AHP andTOPSIS approach was applied to select the optimalalternative for the Harsin earth dam site. To do this, firstly,pair wise comparison matrix of criteria was constructed byusing the experts’ opinions based on saaty’s scale (see Eq. 1).
Then the relative weights of criteria were obtained via Eqs.
2-5, the results are given in Table 2.

205

Table 2 Relative weights of criteria
Criterion
C 1 (+)C 2 (-)
C 3 (+)
C 4 (+)
C 5 (+)
C 6 (+)
C 7 (+)
C 8 (-)
C 9 (+)C 10 (+)
C 11 (-)
C 12 (-)
C 13 (-)
C 14 (-)
C 15 (-)
C 16 (+)
C 17 (+)
C 18 (+)
R

Definition
2T

Health dam site
Overall cost
Annual yield
Topographical conditions
Access to materials and facilities
Economic development
Water quality
Damage of dam body and Reservoir
Volume of reservoir
River flow regime
Water diversion and transfer
Annual volume of sediment
Probable dam break
Probable maximum flood
Average annual evaporation
Environmental impacts
Social impacts
Political impacts

R

R

2T

R

R

R

R

R

R

R

R

R

R

R

R

7T

2T

2T6

2T

2T4

R

R

R

R

R

R

R

R

2T

2T

2T4

R

R

R

R

R

2T4

2T4

2T4

7T

7T

7T

R

R

2T4

2T

R

R

2T6

2T

R

R

2T6

2T

R

R

2T6

2T

2T

2T

Table 4 Normalized weighted fuzzy decision matrix

Relativeweight
0.161
0.155
0.075
0.074
0.045
0.120
0.077
0.053
0.038
0.015
0.046
0.033
0.020
0.020
0.016
0.028
0.017
0.008

Criterion
C 1 (+)C 2 (-)
C 3 (+)
C 4 (+)
C 5 (+)
C 6 (+)
C 7 (+)
C 8 (-)
C 9 (+)C 10 (+)
C 11 (-)
C 12 (-)
C 13 (-)
C 14 (-)
C 15 (-)
C 16 (+)
C 17 (+)
C 18 (+)
R

Criterion
C 1 (+)C 2 (-)
C 3 (+)
C 4 (+)
C 5 (+)
C 6 (+)
C 7 (+)
C 8 (-)
C 9 (+)C 10 (+)
C 11 (-)
C 12 (-)
C 13 (-)
C 14 (-)
C 15 (-)
C 16 (+)
C 17 (+)
C 18 (+)
R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

A
0.000210
0.008606
0.017091
0.000150
0.000090
0.000150
0.000270
0.000150
0.000150
0.005397
0.000150
0.000660
0.000150
0.709743
0.047976
0.000150
0.000090
0.000150

AlternativeB
C
0.000210 0.000030
0.008216 0.008006
0.016612 0.005397
0.000210 0.000270
0.000150 0.000210
0.000150 0.000210
0.000210 0.000150
0.000150 0.000210
0.000150 0.000090
0.004888 0.006957
0.000090 0.000150
0.000690 0.000750
0.000030 0.000270
0.639278 0.226686
0.053973 0.062339
0.000270 0.000150
0.000150 0.000150
0.000210 0.000150

D0.0002700.007196
0.003598
0.000270
0.000270
0.0002100.000150
0.0002100.000090
0.006597
0.000270
0.000810
0.000090
0.151124
0.056971
0.000270
0.000090
0.000150

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

R

In the second stage, the TOPSIS method was applied torank alternatives as follows:
The normalized decision matrix is obtained via Eq. 6.
Obtaining matrix is shown in Table 3. Then, by using Eq. 7,
the normalized weighted fuzzy decision matrix is calculated
and the result is presented in Table 4.
Table 3 Normalized fuzzy decision matrix

A
0.000034
0.001334
0.001282
0.000011
0.000004
0.000018
0.000021
0.000008
0.000006
0.000081
0.000007
0.000022
0.000003
0.014195
0.000768
0.000004
0.000002
0.000001

R

R

R

R

R

R

R

AlternativeB
C
0.000034 0.000005
0.001273 0.001241
0.001246 0.000405
0.000016 0.000020
0.000007 0.000009
0.000018 0.000025
0.000016 0.000012
0.000008 0.000011
0.000006 0.000003
0.000073 0.000104
0.000004 0.000007
0.000023 0.000025
0.000001 0.000005
0.012786 0.004534
0.000864 0.000997
0.000008 0.000004
0.000003 0.000003
0.000002 0.000001

D0.000043
0.001115
0.000270
0.000020
0.000012
0.000025
0.000012
0.000011
0.000003
0.000099
0.000012
0.000027
0.000002
0.003022
0.000912
0.000008
0.000002
0.000001

The PIS (A+) and the NIS (A-) were determined as follows(see Eqs. 8 and 9):A+=(0.0000434, 0.0011154, 0.0012819, 0.0000200,
0.0000121, 0.0000252, 0.0000208, 0.0000079, 0.0000057,
0.0001043, 0.0000041, 0.0000218, 0.0000006, 0.0030225,
0.0007676, 0.0000076, 0.0000025, 0.0000017)
P

P

P

P

P

P

A=(0.0000048, 0.0013339, 0.0002699, 0.0000111,
0.0000040, 0.0000180, 0.0000115, 0.0000111, 0.0000034,
0.0000733, 0.0000124, 0.0000267, 0.0000054, 0.0141949,
0.0009974, 0.0000042, 0.0000015, 0.0000012)
PP

The distance of each alternative from PIS and NIS
solutions was obtained by using Eqs. 10 and 11 as:

d A+ =0.011,

d A− =0.001

d B+ =0.010,

d B− =0.002

dC+ =0.002,

dC− =0.010

d D+ =0.001,

d D− =0.011

Consequently, the relative closeness of the alternative A iwith respect to the PIS can be calculated as shown below (seeEq. 12):
R

R

0.001
= 0.0850.011 + 0.001
0.002
CC B == 0.150
0.010 + 0.002
0.010
CCC =
= 0.8450.002 + 0.010
CC A =

CCD =

0.011
= 0.9160.001 + 0.011

The relative closeness to the PIS clearly shows the ratingof alternatives. Based on these values (CC i ), four proposed
R

206

R

alternatives were ranked “D”, “C”, “B” and “A”, respectively.
This means that, the alternative ‘D’ is the optimal alternative
for the Harsin earth dam site. Fig. 3 shows the Front view ofthis alternative.

Fig. 3 Front view of the optimal site (alternative D)
5 CONCLUSION
The dam site selection is related to many factors.
Therefore, multi-criteria decision making could be used tosolve this problem. In this study, AHP and TOPSIS, which
are two of the most powerful and flexible methods for use insolving multiple criteria problems were used in an integrated
approach for locating the optimal earth dam site, thisapproach was applied to locate the Harsin earth dam site at thewest of Iran. For this purpose, four alternatives A, B, C and D,
were evaluated with regard to the most effective criteria.
After applying this integrated approach, alternative D was
identified as the optimal site for the Harsin earth dam. The
Merit of this site (Alternative D) was also approved byexperts involved in this project. Therfore, this approach could
be used in selecting the optimal site in other dams.
6 ACKNOWLEDGMENT
The authors would like to appreciate Kermanshah
Regional Water Company and Pandam Consulting Engineers
Company for providing required information and data.
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20, 1978, pp. 3–24.
[7] Saaty TL, The analytic hierarchy process. New York:
McGraw-Hill, 1980.
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[9] Brans JP, Vincke P, Mareschal B, ”How to select andhow to rank projects: the PROMETHEE method”, Eur.
J. Oper. Res., vol. 24, 1986, pp. 228-238.
[10] Deng JL, “Introduction to grey system theory”, The
Journal of Grey System, vol. 1, 1989, pp. 1-24.
[11] Janssen R, Multiobjective Decision Support forEnvironmental Management. Netherlands: Kluwer
Academic Publishers, 1996.
[12] Opricovic S, Multi-criteria optimization of civilengineering systems. Belgrade: Faculty of Civil
Engineering, 1998.
[13] Opricovic S, Tzeng GH, “Multicriteria planning of post
earthquake sustainable reconstruction”, Comput-Aided.
Civil. Infrastruct. Eng., vol. 17, 2002, pp. 211–220.
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alternatives using ELECTRE”, Hydrolog. Sci. J., vol.
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208

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Using Geogrid Boxes as a New Approach for Reinforcement of Rock Slopes
Ahmad Fahimifar a, Arvin Abdolmaleki baDepartment of Civil Engineering, Amirkabir University of Technology, Tehran, Iran.
b

Department of Civil Engineering, Tafresh University, Tafresh, Iran.

ABSTRACTUse of geosynthetic reinforcement in structures such as soil slopes, paved or unpaved roads and runways, has been thesubject of wealth of research works. However, there is no published work that encompasses the scope of stabilizationof rock slopes using geogrid boxes. The main objective of this research is to introduce the geogrid box reinforcement
method as a new approach to reinforce rock slopes and investigate whether the new approach can be a satisfactory
alternative for the other conventional slope stabilization methods. The limit equilibrium (LE) and finite element (FE)analysis were conducted to predict the response of the slope to a broad range of possible conditions, namely dry, halfsaturated and saturated states; as well as, static, quasi-static and dynamic conditions. Results confirmed that installing
geogrid box can be used as a satisfactory alternative for the rock bolt system method.
Keywords: Rock slope stability, geogrid box, rock bolt, limit equilibrium analysis, finite element analysis.

1. INTRODUCTION
Geosynthetics have become well established construction
materials for geotechnical and environmental applications.
In the last two decades the use of geosynthetic materials forreinforcing slopes and retaining walls in fill has increased
extensively throughout the world. In slope and wall
applications the function of the geosynthetic layers is tooffer resistance to driving forces or moments caused by the
self-weight of the soil and applied surcharges [1], [2].
Wealth of research have been carried out on reinforcementof soil slopes by geosynthetics, most of which highlight theadvantages of using this method over other reinforcement
methods [1]-[3]. San and his associates [1] indicated that FE
and LE methods could consistently determine the locationsof critical slip surfaces and required tensile strength ofreinforcement in geosynthetic-reinforced slopes. Despiteextensive research on geogrid-reinforced soil slopes, weknow of hardly any published work that covers the scope ofstabilization of rock slopes using geogrid boxes. Therefore
our objectives at the present work are to introduce thismethod as a new approach to reinforce rock slopes andcompare this method with other conventional slope
stabilization methods.
2. EXPERIMENTAL
STUDY

PROCEDURE

AND

CASE

The study area is located approximately along km 11+060 of
Mianeh-Ardabil railway track. This section is problematicdue to the existence of lithological units with variablecharacter. The Engineering geological properties of therocks exposed in the case study slopes and technicalspecifications of geogrid are presented in Table I. Fig. 1
depicts the profile of site.
Three design methodologies were presented to investigate
the effect of height and width of geogrid boxes on the factorof safety. The factor of safety for slope stability analysis isdefined as the ratio of the ultimate shear strength divided by
the mobilized shear stress at incipient failure.

209

Table I. Required parameters for stability analysis and geogrid parameters

Parameter

SymbolUnit

CohesionC

kPa

Internal friction angle

Unit weight

Elasticity modulus

Poisson’s ratio

Geogridparameters

Horizontal earthquake
acceleration in quasi-static
analysis
Tensile Strength
Cohesion
Friction angle
Equivalent axial stiffnessϕγRock type

degree

KN/m3
P

E

ν

KN/m2
P

--

R-I
R-II
R-III
R-used in geogrid
box
R-I
R-II
R-III
R-used in geogrid
box
R-I
R-II
R-III
R-used in geogrid
box
R-I
R-II
R-III
R-used in geogrid
box
R-I
R-II
R-III
R-used in geogrid
box

Value

570
50
390
140
60
25
55
40

26.5
24.5
25
26.5
6.067 e + 09
3.77 e + 07
2.42 e + 09
6.06 e + 08
0.25
0.32
0.27
0.25

Kh

m/s2

--

0.1g

NP
C
ϕ
EA

KN/m
KN/m2
degree
KN/m

--

200
5
40
1.71 e + 04

R

R

210

P

P

---

Fig. 1 Profile of the slope
decreased the factor of safety. For the LE analysis we
adopted the Spencer method which assumes a constant
3.1. Stability analysis of the excavated slope
relationship between the magnitude of the interslice shear
and normal forces [4]. This method satisfies both the
The safety factors obtained for the excavated slope in moment equilibrium and force equilibrium. The only
different conditions are shown in Table II. The results limitation of this method is the need to a software programindicate that excavation of the slope has considerably to conduct the complicated calculations.
Table II. Factor of safety for excavated slope in Plaxis and Slide
3. RESULTS AND DISCUSSIONS

Dry
Half-saturated
Saturated

Static
1.13
1.19
1.08

Factor of safety(FE based software)
Quasi-static
Dynamic
1.00
1.05
0.94
0.98
0.91
0.94

The different safety factors obtained can be attributed to theanalysis mode performed by PLAXIS and SLIDE. PLAXIS
performs numerical analysis in which no pre-defined slip
surface or inter-slice forces are required. Furthermore, thepossible slip in FE method takes place in a zone. On theother hand, in LE based software SLIDE, the shape andposition of slip surface have to be priori defined and theslide surface is considered to be a narrow curve. Considering
above, the outputs of numerical methods seem to be more
valid than that of limit equilibrium methods [5]-[10].

Factor of safety(LE based software)
StaticQuasi-static
1.28
1.13
1.24
1.07
1.07
0.94

The calculations were carried out for the different conditions
with a maximum height of 8 m for each bench by selecting athickness of 1 m for every reinforcement layer. At the
beginning of each bench the width of geogrid box is 10 m.
The width of the upper box is 9.5 m and the 0.5 m reduction
in box width will be continued so that the box width at theend of each bench reaches 6.5 m. Fig. 2 depicts theschematic of Plan 1. The results of stability analysis arepresented in Table III. As is observed, the factor of safety
has increased to a considerable extent.

3.2. Plan 1

211

Fig. 2. Schematic of geogrid boxes installation in plan 1

Table III. Factor of safety for plan 1 in Plaxis and Slide

Dry
Half-saturated
Saturated

Static
1.70
1.64
1.40

Factor of safety(FE based software)
Quasi-static
Dynamic
1.44
1.52
1.39
1.48
1.30
1.39

3.3. Plan 2
In this plan, at the beginning of each bench the width ofgeogrid box is 15 m. The width of the upper box is 14.5 m
and the 0.5 m reduction in box width will be continued sothat the box width at the end of each bench reaches 11.5 m.
All the boxes are 1 m in height.

Factor of safety(LE based software)
StaticQuasi-static
1.75
1.53
1.72
1.50
1.66
1.37

The results of stability analysis are presented in Table IV. As
is observed, in comparison with plan 1 the factor of safety
decreases. Further increase in box width requires larger
amount of excavation, as a result of which, the incline ofupper eroded slope increases, and hence the factor of safety
decreases.

Table IV. Factor of safety for plan 2 in Plaxis and Slide

Dry
Half-saturated
Saturated

Static
1.42
1.41
1.40

Factor of safety(FE based software)
Quasi-static
Dynamic
1.22
1.31
1.20
1.28
1.18
1.27

3.3. Plan 3
In this plan at the beginning of each bench the width ofgeogrid box is 5 m. The width of the upper box is 4.5 m andthe 0.5 m reduction in box width will be continued so that

Factor of safety(LE based software)
StaticQuasi-static
1.66
1.44
1.63
1.41
1.54
1.32

the box width at the end of each bench reaches 1.5 m. All theboxes are 1 m in height.
The results of stability analysis are presented in Table V. As
is observed, the factors of safety obtained are lower thanthose obtained in plans 2 and 3. Referring to Tables III-V

212

and regarding the amount of consumed geogrid, plan 1 isselected for reinforcement of the rock slope and comparison

with rock bolt system method.

Table V. Factor of safety for plan 3 in Plaxis and Slide

Dry
Half-saturated
Saturated

Static
1.40
1.36
1.29

Factor of safety(FE based software)
Quasi-static
Dynamic
1.15
1.24
1.13
1.22
0.85
0.94

3.4. Rock bolt system
Based upon available information, the ribbed bolts were
installed using 3 × 3 staggered pattern [11]. A hex nut andbearing plate are used to distribute the tensile load from thebolt to the rock mass. Both surface and subsurface drains
were installed to decrease the pore pressure acts on the rockmass and hence increasing the factor of safety. Subsequentstabilization measures were also implemented including
adding shotcrete. To increase tensile strength and structural
support, the shotcrete was used with welded wire mesh.

Factor of safety(LE based software)
StaticQuasi-static
1.49
1.12
1.48
1.11
1.42
1.09

The results of stability analysis using rock bolt system areshown in Table VI. Considering above plan 1 enjoys the
maximum factor of safety. In comparison with rock bolt
system, implementing plan 1 causes cost reduction due todecrease in amount of excavation, allowing the slope tostand at steeper angles, reduction in labor, increasing thefactor of safety, omitting some of the machinery andutilizing the existence material in the place, as well as
reduction in construction time. This method also can besatisfactorily used to meet a variety of geological conditionsand various support requirements.

Table VI. Factor of safety for rock bolt system in PLAXIS and SLIDEDry
Half-saturated
Saturated

Static
1.38
1.32
1.23

Factor of safety(FE based software)
Quasi-static
Dynamic
1.27
1.31
1.20
1.23
1.12
1.12

4. CONCLUSION
Stability analyses are routinely performed in order to assess
the safe and functional design of an excavated slope. The
analysis technique chosen depends on both site conditions
and the potential mode of failure, with careful consideration
being given to the varying strengths, weaknesses andlimitations inherent in each methodology. The limit
equilibrium (LE) and finite element (FE) analysis wereconducted to predict the response of rock slopes to a broad
range of possible scenarios, namely dry, half-saturated andsaturated states; as well as, static, quasi-static and dynamic

Factor of safety(LE based software)
StaticQuasi-static
1.40
1.33
1.31
1.25
1.12
1.05

conditions. The results indicate that the factors of safetyobtained through the geogrid box method are higher thanthose obtained through rock bolt system method.
5. REFERENCES
[1]

[2]

213

San, K.C., D. Leshchinsky, and T. Matsui,"Geosynthetic Reinforced Slopes: limit equilibrium
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Ahn, T.B., S.D. Cho, and S.C. Yang, "Stabilizationof Soil Slope Using Geosynthetic Mulching Mat,"
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Stabilization, in Stability and Performance ofSlopes and Embankments1994: Berkeley
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Mahboubi, A., M. Aminpour, and A. Noorzad,
Conventional and Advanced Numerical Methods ofRock Slope Stability Analysis, A Comparison
Study, Gotvand Dam Right Abutment (Iran) Case

[8]

[9]

[10]

[11]

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Study, in the "12th International Conference ofInternational Association for Computer Methods
and Advances in Geomechanics (IACMAG)2008"
Goa, India
Huang, M. and C.Q. Jia, "Strength Reduction FEM
in Stability Analysis of Soil Slopes Subjected toTransient Unsaturated Seepage," Computers andGeotechnics, 2003. 36: p. 93-101.
Gurocak, Z., S. Alemdag, and M.M. Zaman, "RockSlope Stability and Excavatability Assessment ofRocks at the Kapikaya Dam Site, Turkey,"
Engineering Geology, 2008. 96: p. 17-27.
Alkasawneh, W., et al., "A Comparative Study of
Various Commercially Available Programs in
Slope Stability Analysis," Computers andGeotechnics, 2008. 35: p. 428-435.
Abramson, L.W., et al., Slope Stability andStabilization Methods. 2 ed. 2002: John Wiley andSons.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Volume Change Characteristics of Bentonite Kaolinite Mixes
J. M. Kate
Datta Meghe Institute ofEngineering, Technology & Research,
Wardha - 442004, India

Sunil Kumar
Former M. Tech student,
Indian Institute of Technology
Delhi, New Delhi- 110016, India

M. P. Bhorkar
Datta Meghe Institute ofEngineering, Technology & Research,
Wardha - 442004, India

ABSTRACTThe fat clay commercially known as ‘Bentonite’ has extensive utilities in construction industries. The present
experimental study has been undertaken to understand swelling pressure and volume change behavior of bentonite andkaolinite (kaoline clay) mixed in different proportions by weight. Mathematical equations correlating bentonite
percentages with swelling, swelling pressure and related index properties have been developed. The studies clearly
indicate considerable influence of bentonite on volume change behavior and related properties of clay mixes. The
findings of the present study and the recommendations given may prove to be useful in construction industry where
bentonite is extensively used as a single material or mixed with other clays.
Keywords: Bentonite, Free Swell Index, Kaolinite, Swelling Pressure, Volume Change

1. INTRODUCTION

2. LITERATURE REVIEWExtremely fine grained fat clay commercially known as‘Bentonite’ is composed of montmorillonite as its
predominant mineral constituent. The properties such as very
high volume change (swelling and shrinking), swellingpressure, high cohesion, extremely low permeability,
thixotrophy, etc. possessed by bentonite clays make it
extremely useful material in construction industry especiallyfor specific civil engineering construction, exploration andground improvement practices. Just to list a few of utilities, it
is extensively used during the construction of diaphragm wall,
also used in slurry forms as bore hole stabilizer , fine clay
grout, drilling mud, etc. Processed bentonite clay is invariably
adopted as a sealing / lining material for industrial waste
repositories and landfills to prevent percolation of liquid
pollutants in to the surroundings. It is used as a single material
only or in combination with other clays mostly on economy
considerations. Processed kaoline clay having kaolinite as its
predominant mineral constituent, is chosen for thisinvestigation as it possesses negligible swelling and related
properties.
In view of numerous utilities and related specification
requirements, the present experimental study has been
conducted to understand the volume change, swellingpressure behavior of clay mixes of bentonite and kaoline clay
in different proportions. Here in an attempt has been made to
develop mathematical equations to correlate bentonite
percentages with volume change (swelling), swellingpressure. The findings have been discussed in the light ofindustrial applications of this study. The recommendations
pertaining to suitability of clay mixes for particular job havebeen given.

2.1 Overview
The swelling phenomenon in clays is attributed mainly to thepresence of montmorillonite clay mineral in them. Bentonite
clay contains sodium montmorillonite as its predominant
mineral constituent. The governing property of bentonite tosatisfy specification requirements is different in different
utility jobs. However, all such properties are directly relatedwith degree of expansivity & swelling potential, e.g. in its use
as a sealing material in underground subsurface waste
repositories, the governing property/ criterion is its extremely
low permeability. The clays with higher degree of expansivity
exhibit lower permeability [1]. Similarly all other governing
properties/ criteria can be assessed/ predicted from swelling/
volume change behaviour of expansive clays.
2.2 Engineering behaviour of expansive clays
Experimental investigation was carried out by Kate [1] on
two extreme (low and high) grades of Indian bentonites toassess their suitabilities for utilities in civil engineering
construction works. The relevant properties determined weremineralogy, chemical compositions, swelling, swellingpressure and coefficient of permeability. The bentonite slurry
behaviour was assessed through viscosity, get strength, pH,
sedimentation and electrical conductivity. He reported thateven without any additive both these extreme grade
bentonites satisfy minimum specification requirements fortheir uses in almost all Civil engineering construction works.
Based on synthesis of data available in literature Sridharanand Prakash [2] summarized broad guidelines to predict
degree of expansivity of clays from routine tests. Accordingly,
from the magnitudes of liquid limit, plasticity index, colloidal
clay contents, shrinkage limit, free swell index, etc. the215degree of expansivity (low, medium, high or very high) canbe predicted. The synthesis and analysis of available datafrom literature on swelling, swelling pressure and otherrelevant engineering characteristics of expansive clays werecarried out by Sunil Kumar [3]. He reported good correlations
between free swell index, maximum percentage swell andcorresponding swelling pressure of such clays.
3. EXPERIMENTAL PROGRAMME
3.1 Clays & clay mixes
Commercially available processed bentonite clay and kaoline
clay (Kaolinite) passing through 425-micron sieve have beenused for the present study. Representative clay mixes wereprepared by thoroughly mixing absolutely dry (zero moisture
content) bentonite and kaolinite in desired proportions byweight. The proportions of bentonite and kaolinite adopted
were 100:0, 80:20, 70:30, 50:50 and 30:70, which for brevity
are referred in the text as clay mix A, B, C, D and E
respectively.
3.2 Tests Conducted
Both the clays (bentonite, kaolinite) and clay mixes have beentested for their basic physical and engineering properties
following the procedure as per Indian standard codes ofpractice for various laboratory tests. These tests include
specific gravity, specific surface, hydrometer analysis andStandard Proctor test (for maximum dry unit weight and
OMC). Chemical analysis of bentonite and kaolinite has also
been carried out. Atterberg’s limits such as liquid limit,
plastic limit, plasticity index and shrinkage limit of all theseclays & mixes were determined. Free Swell Index tests onclay mixes were conducted adopting the procedure suggested
in IS: 2720, part 40 [4].
The maximum percentage swell and swelling pressure were
determined as per the procedure given in International
standard TC-6 [5]. Accordingly, the maximum percentage
swell is the ratio of heave (under applied stress of 5kPa) to theinitial thickness of clay sample in Oedometer. The percent
swell has been obtained under each applied stress followed bydetermination of swelling pressure (which corresponds tomagnitude of applied stress at zero percent swell) from theplot of percent swell versus applied stress (on log 10 scale).
Swell and swelling pressure tests have been conducted on allthese clay samples remoulded at initial compaction unit
weight of 12.6 kN/m3, which corresponds to Standard Proctor
maximum dry unit weight of bentonite. The initial moisture
content for all clay mixes was maintained at zero percent as itprovides maximum value of swelling pressure corresponding
to chosen compaction dry unit weight.
The procedure adopted to compact the samples was that,
calculated quantity (weight) of clay mix corresponding toinitial unit weight of 12.6 kN/m3 and zero moisture content
was placed in Oedometer ring attached with collar. This mix

was then compressed upto desired thickness in the ring using
plunger plate (of same diameter as ring) and hydraulic jack.
These clay samples in the ring where subjected to full water
saturation under desired applied stresses during whichchanges in their thicknesses were recorded.
4. RESULTS AND DISCUSSION
4.1 Engineering properties and chemical composition
The physical and engineering properties of clays and theirmixes are illustrated in Table I.
Table I

Physical and Engineering properties of clays

A
2.71

Clay Samples
BC
D
E Kaolinite
2.71 2.70 2.68 2.65 2.63

Specificsurface (m2/g)

135-

Maximum dry
unit weight(kN/m3)

12.6

OMC (%)

Property
Specific
gravity

-

-

-

14

13.2 14.0 14.9

15.1

16.9

41

26

24

23

21

20

Silt (%)(0.075mm0.002mm)

22

26

28

35

39

46

Clay (%)(<0.002mm)

78

74

72

65

61

54

Colloidal clay(%)(<0.001mm)
Classification

47

41

37

35

31

24

CH

CH

CH

CH

CH

CH

A-100*:0+ B-80*:20+ C-70*:30+ D-50*:50+ E-30*:70+*Bentonite Clay (%), + Kaolinite Clay (%)
The Table clearly shows distinct trend of changes in all theproperties brought due to decreasing percentage of bentonite.
For example, the maximum dry unit weight continuously
increases from 12.6 kN/m3 for 100% bentonite (Sample A) to
16.9 for 0% bentonite (100% Kaolinite). Such trends ofeither continuous decrease or increase in magnitudes ofproperty with decrease in bentonite percentage are distinctly
noticeable.
The percentages of colloidal clay (<0.001 mm) in these mixes>27% indicate that, they possess very high degree of
expansivity [2]. All the clays and clay mixes are classified(‘A’ line classification) as CH i.e. clays with high plasticity.
The chemical composition of bentonite and kaolinite ispresented in Table II, which shows SiO 2 is their predominantconstituent followed by Al 2 O 3 .

216

Table II

Chemical composition of clays

ChemicalComposition (%)
SiO 2
Al 2 O 3
Fe 2 O 3
CaO
MgO
Na 2 O
K2 O
SO 3
OthersLoss on Ignition

Clay
Bentonite
60.23
15.16
8.39
4.78
2.18
2.57
0.83
0.25
1.23
4.38

Kaolinite
53.06
28.98
2.03
0.84
0.12
0.17
2.66
0.11
2.97
9.06

Fig. 2. FSI and SL as a function of bentonite percentage

4.2 Atterberg’s limits and Index properties
The variation of liquid limit (LL), plastic limit (PL) andplasticity index (PI) with bentonite percentages present inclay mixes is illustrated in Fig. 1. It shows all increases
exponentially with increase in bentonite percentages.

It is seen from this figure that with increase in bentonite
percentage the FSI increases exponentially, whereas SL
decreases. The values of FSI (>200%) exhibited by clay
mixes with bentonite percentage of 70 and above fall in very
high expansivity, whereas the remaining clay mixes belongs
only to high expansivity category as per [2].
4.3 Maximum swell and swelling pressure
The plots of swell versus applied stress for these clay mixes
are shown in Fig. 3. The swell decrease with increasing
applied stress as represented by variational curves in Fig. 3.
The magnitudes of swelling pressure have been obtained
from these curves.

Fig. 1. Variation of Atterberg limits with bentonite
percentages
The magnitudes of LL (>60%) and PI (>32%) exhibited byclay mixes demonstrate very high degree of their expansivity[2]. The regression analysis of the data provides thecorrelations expressed by Equations (1), (2) and (3)0.022(B)

LL = 52.74e

2

(1)

2

(R = 0.988)

2PL = 0.0033(B) + 0.16 (B) +12.14(R = 0.915)(2)

PI = 0.783(B)1.331(R2 = 0.878)(3)

Wherein, B denotes bentonite percentage and R2 is coefficientof correlation
Free Swell Index (FSI) and shrinkage limit (SL) as a functionof bentonite percentage in these clay mixes is shown in Fig. 2.

Fig. 3. Effect of applied stress on swell for clay mixes
The swelling pressure exhibited by clay mix A, B, C, D and E
are 425, 340, 255, 175 and 106 kPa respectively and thecorresponding maximum swells are 21.86, 17.02, 14.45, 9.00
and 4.83 percents. These values of swelling pressures as wellas maximum swell (%) fall under very high for A, B & C andhigh/ moderate categories for D & E.
The swelling pressure as well as maximum swell as a functionof bentonite percentage demonstrates that both increases withincrease in bentonite as illustrated by curves in Fig. 4.

217

It is seen in Fig. 5 that both the curves exhibit nearly the same
trend of power law relationships. It is interesting to note thatthese curves show significant deviations at lower values ofswelling pressures, which tends to reduce with increasing
values and beyond swelling pressure of 250 kPa both thesecurves overlap. This reflects on the distinct behaviour of
bentonite and its mixes as compared to other expansive clays.
The synthesized data from the literature corresponds mostly
to clays having combination of montmorillonite and illite as
their mineral constituents.
5. CONCLUSIONS
Fig. 4. Swelling pressure and maximum swell as a function of
bentonite percentage
The relationship between swelling pressure (P sw ) andbentonite percentage is expressed by Eq. (4).
P sw = 0.018(B)2 + 2.37(B) + 16.06(R2 = 0.855)(4)

4.4 FSI and Swelling Pressure
A synthesis of available data on FSI and corresponding
swelling pressure for various expansive clays extracted frommost recent literature has been carried out in the present study.
Various points on the plot of FSI versus swelling pressure
shown in Fig. 5 correspond to such data from several
investigators [3].

The findings of the present experimental study lead to thefollowing conclusions .
(i) The degree of expansivity as predicted from semi-direct
criteria such as colloidal clay contents and Atterberg’s limits,
most of the clay mixes fall in the category of very highexpansivity.
(ii) The prediction through direct criteria i.e. FSI, percent
swell and swelling pressure, the clay mixes with 70% andabove bentonite fall in very high expansivity category & theremaining mixes (50% and 30% bentonite) between high tomoderate category.
(iii) The synthesized data from literature and present
experimental data demonstrate a power law relationshipbetween FSI and swelling pressure covering a very wide
range of their magnitudes.
6. RECOMMENDATIONS
The clay mix (with comparable quality of bentonite andkaolinite) having 70% and above bentonite is reccomended
for its use in construction of diaphragm wall, sealing/ lining
for industrial waste respositories, etc. Further, it isreccomended to use clay mixes having 50% or only 30%
bentonite for the work where used in slurry form e.g. fine clay
grouts, borehole stabilizer, drilling mud, etc.
7. REFERENCES

Fig. 5. Synthesized and experimental data of swellingpressure versus FSI
The curve in this figure has been derived as a best fit
representing the entire data, and its regression analysis
provides the correlation expressed by Eq. (5).
P sw = 4.46 (FSI)0.725(5)(R2 = 0.921)

The experimental results of the above expansive properties
obtained in this study are also illustrated in Fig. 5. The
experimental values of swelling pressure of these clay mixes
correlates with FSI as given in Eq. (6).
P sw = 4.1(FSI)0.76(6)(R2 = 0.953)[1] Kate J M, “Mineralogical and Engineering Behaviour of
Indian Bentonites”, Proc. 2nd World Congress on
Non-metallic Minerals, Beijing, China, Oct. 1989, pp.
393-398.
[2] Sridharan A and Praksh K, “Classification Procedures for
expansive Soils”, J. of Institution of Civil Engineers,
Geotechnical Engineering, vol. 143, 2000, pp. 235-239.
[3] Sunil Kumar, “An Experimental Study on the Behaviourof Expansive Clays treated with different Fly Ashes”,
Unpublished M. Tech. Thesis CE Deptt., IIT Delhi, May
2007, pp.1-158.
[4] Indian Standard IS: 2720, Part 40, “Determination of Free
Swelling Index of Soils”, Bureau of Indian Standards,
New Delhi, 1997, pp. 1-17.
[5] Technical Committee TC- 6, “Evaluation of SwellingPressure and corresponding Heave of Expansive Soils byconstructing Swell percentage versus applied total stress218

diagram”, Report of Technical Committee (ISSMGE) on
expansive soils, 1993, CBIP, New Delhi, pp. 1-26.

219

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

FE Analysis for Settlement of Peat Ground under Embankment Loading

Satoshi Nishimoto and Hirochika Hayashi
Civil Engineering Research Institute for Cold Region, Japan
ABSTRACT
Peaty ground commonly found in Hokkaido, Japan, is an extremely soft ground, which is highly organic and has special
engineering properties. Finite element analysis (FE analysis) with consideration of the overlay pavement load as ameasure against residual settlement was thus conducted concerning the long-term settlement of a road on peaty soft
ground, and the relationship between the residual settlement and repair cost was studied. It was found that FE analysis
using the Sekiguchi-Ohta model can express the long-term settlement of peaty soft ground. It was also revealed that theamount of residual settlement greatly affects the repair cost.
Keywords: peat, FE analysis, monitoring, settlement, life cycle cost, expressway embankment

1. INTRODUCTION
Peaty ground distributed widely in Hokkaido, Japan, is anextremely soft ground, which is highly organic and has
special engineering properties. Roads on peaty ground areusually designed to allow for a certain degree of settlement
and are repaired and maintained during their service, sinceconsiderable settlement occurs over a long period of time insuch roads.
The Manual for Peaty Soft Ground
Countermeasures [1] stipulates the allowable residual
settlement as between 10 to 30 cm in three years after an
expressway is placed in service. These values were, however,determined empirically and should be reconsidered for theminimization of the life cycle cost (LCC).
FE analysis with consideration to the overlay pavement
load as a measure against residual settlement was thus
conducted concerning the long-term settlement of a road on
peaty soft ground, and the relationship between the residual
settlement and repair cost was studied.

Figure 1. A typical settlement curve of peaty ground

2. SETTLEMENT BEHAVIOR OF PEATY GROUND
AND THE CONCEPT OF THE LIFE CYCLE COST
Figure 1 is a conceptual diagram of the settlement of peaty
ground. A characteristic of the settlement of peaty ground is
the long-term settlement called secondary consolidation,
which occurs linearly to the logarithm of time and causes
continuous settlement after the placement of roads in service.
The type of ground, where soft clay layers often accumulate
at the bottom of peat, is called peaty soft ground. In addition
to the peculiarity of peat, the existence of thick clay layers
makes the problem of long-term settlement even more
complex.
The standard method in the Peaty Soft Ground219

Figure 2. The concept of the life cycle cost

Figure 3. The cross section of the embankment
Countermeasures Manual for the construction of a roadembankment on peaty soft ground is the method of allowing
for a certain degree of residual settlement after the placementof the road in service and conducting maintenance and repair
works during its service life. This method is used for
reducing road construction costs, as well as for ensuring theflatness of the road surface and con-trolling road repair costs
depending on the importance level of the road. It means thatthe concept of LCC shown in Figure 2 has been introduced.
There are, however, some problems that must be solved forrationalization, such as the empirical setting of allowable
residual settlement.
3. OVERVIEW OF FE ANALYSIS
3.1 Overview of the section of analyzed
The subject of the analysis, Mihara Bypass (1.4 km long), is
an expressway on the outskirts of Sapporo, which is thelargest city in Hokkaido. It is a two-lane road constructed on
peaty soft ground. To ensure stability and reduce residual
settlement, the preload method was adopted in combination
with the counterweight fill method, which is the use ofloading berms at the sides of the main embankment (Figure 3).
Measurements of settlement have been continued since thisexpressway was placed in service in March 2005.
3.2 Analysis conditions
FE analysis was conducted using the Sekiguchi-Ohta model[2], which can express long-term settlement. Table 1 presents
the analysis cases. A comparison between measured
settlement and analysis results was first made as a preliminary
study, and the applicability of the constitutive model andvalidity of parameters were verified.
Next, in the main study, virtual simulation was
conducted for three cases with different amounts of residual
settlement three years after their placement in service. At thattime, the ground and other conditions were assumed to beuniform and the specific amounts of settlement were set byvarying the period of waiting time for decreasing settlement.
Concerning the bumps caused by settlement, it was assumed
that repairs (overlay pavement) were repeated whenever220

Table 1. Analysis cases and conditions
analysis conditions
caseoverlay pavement
load

residualsettlement (cm)preliminary
case

no loading－

1

loading302loading20

3loading

10

main
case

settlement reached 10 cm, based on the results of thefact-finding survey of repairs conducted by Nishimoto andHayashi [3]. As the loading condition at that time, the load of
the pavement thickness of 10 cm was uniformly distributed
over the embankment crest, thus causing a slight increase inload from the embankment at this point. The traffic load
generated by vehicles passing over the bumps was not taken
into account because the embankment was high.
3.3 Determination of soil parameters
When analyzing the long-term settlement of peat, it isimportant to determine the coefficient of permeability andcoefficient of secondary consolidation. Hayashi et al. [4]have revealed that the coefficient of permeability of peat
decreases considerably with consolidation and that the actual
coefficient of permeability is 10 to 30 times as large as thevalues found by the oedometer test. These are engineering
properties peculiar to peat. In this analysis, therefore, theinitial coefficient of permeability was determined by Eqs. (1)and (2). The changes in the coefficient of permeability
associated with consolidation were given by Eq. (3). Where,
koed is the coefficient of permeability found by the oedometer
test (cm/s), Cv is the coefficient of consolidation (cm2/day),
mv is the coefficient of volume compressibility (m2/kN), w is
the unit weight of water (= 9.81kN/m3), k0 is the initialcoefficient of permeability (cm/s), koed-p0 is the coefficient of
permeability at the time of effective overburden pressureconvenient method

detailed methodsoil parameters②M①

ignition loss L i③

consolidated-undrained
triaxial compression

'

K 0NC

④

triaxial
K0-consolidation test

m

 V0'

OCR

⑥

⑦

⑧

Pc

K 0OC
⑨

v
⑩

k f/k lkl
k const.

tnatural water content
Wn

wet density test

⑤

⑬

⑪


⑭



⑫

oedometer test⑮

soil particle densityse0①  ' = 0.19Li (%) +32⑥  V0' =  th -P w

② M = 6sin ' /(3-sin ')Jaky (1948)③ K 0NC = 1-sin '⑦ OCR = P c / V0'⑧ K 0OC = K 0NC OCR⑨ v = k 0 /(1+k 0 )④ m = 0.005Li (%) +0.45⑤  t = (( s- w)/(1+e ))g n⑪  = C c/2.3⑫  = C s/2.3⑬ C c = 0.01W n(%)
⑭ C s = 0.1C cm⑮ e 0 =  s/ w W n(%)⑩ k f/k l = 0.33L i (%)

Figure 4. Determination procedures of soil parameters for FE analysis of peat and organic clay (Hayashi et al, 2007b)
Table 2. The list of soil parameters usedsoil layer

coefficient of irreversibility critical state compression
dilatancy Dparameter M
ratio index initial void
ratio e 0

coefficient of
secondaryconsolidation

initialvolumetricstrain rate
V 0 (day/1)
2.2×10coefficient of permeability
horizontalk H (m/day)

0.21

0.30

0.26

-5

0.30

0.43

0.43

2.6×10

-6

0.37

0.78

0.58

8.6×10

-5

0.35

0.71

0.54

1.6×10

-6

0.36

0.59

0.57

9.9×10

0.10

0.92

1.94

2.69

12.36

0.03

organic clay

0.15

0.90

1.48

1.62

5.44

0.02

3.3×10

clay

0.11

0.80

1.08

0.39

1.63

0.01

2.9×10

silt

0.07

0.82

1.16

0.20

0.91

0.01

2.3×10

clay

0.09

0.86

1.10

0.24

1.06

0.01

1.4×10221

coefficient of coefficient ofearth
in-situ earthpressurepressureat rest K 0
at rest K 0OC

-5

peat

found by the oedometer test (cm/s), e is the void ratio at theend of consolidation and e0 is the initial void ratio. The
constant Ck in Eq. (3) was determined from the relationshipbetween the void ratio and the coefficient of permeability

Pisson's
ratio v

-2

Verticalk V (m/day)change withvoid ratio k

1.7×10

-2

1.16

-2

5.2×10

-3

0.84

-4

8.6×10

-4

0.43

-3

1.6×10

-3

0.24

-4

9.9×10-4

0.31

8.6×10found by the oedometer test.
koed = Cv mv w / (8.64×106)(1)

k = k0 exp((e-e0) /Ck)

(3)

12

thickness of fill
(m)

(2)

Noto [5] related the coefficient of the secondary
consolidation Cs (%) of peat with the natural water content
Wn and expressed the relation by Eq. (4). The coefficient of
secondary consolitaion in this analysis was determinedusing Eq. (4).
(4)

Other soil parameters of peat were determined fromthe results of laboratory soil tests using the flow chart(Figure 4) proposed by Hayashi et al. [6]. The parametersof clay were deter-mined using the method of Iizuka andOhta [7]. Table 2 lists the soil parameters used for analysis.

settlement
settlement
(cm)
(cm)

Cs (%) = 3.3+0.0043Wn (%)

8
prototype4

analysis

0
0

1000

2000

3000

0
0

elapsed time (days)
1000
2000
1000
2000

3000
3000

0
0

4. RESULTS OF PRELIMINARY STUDY
Figure 5 displays the changes in measured and analyzed
ground surface settlement at the center of the embankment
with the passage of time. The embankment constructed at
the section was 10.8 m in thickness, and the measured
settlement at the start of service was approximately 4 m,
indicating the high compressibility of the ground. Measured
values have been obtained for 2,075 days (684 days inservice) so far, and almost correspond with the analysis
results.
This means that FE analysis using the
Sekiguchi-Ohta model is effective in the examination of thelong-term settlement of peaty ground. It also shows that theprocedure to determine soil parameter presented in thispaper was valid.

measuredmeasured
analyzed
analyzed

100
100
200
200
300
300
400
400
500
500
600
600

Figure 5. The changes in measured and analyzed groundsurface settlement at the center of the embankment
elapsed time (days)
0

5000

10000

15000

20000

25000

0
measured100
settlement (cm)k0 = 10 koed-P0200

analyzed (fill load)
start of service

analyzed (fill + overlay)

300
50 years
400
500
× : overlay pavement
600

Figure 6 presents the analysis results of Case 1 (30cm
residual settlement in three years after placement in service).
According to the analysis results, settlement will continue
after placement in service and the residual settlement in 50
years is expected to be approximately 92 cm only in thecase with the embankment load and 114 cm in the casewhere the overlay load is also taken into account. While theoverlay load is rarely considered in normal settlement
analysis, it is worth noting that, in the analysis results of thisstudy, the amount of residual settlement was 22 cm greaterwith the overlay load than in the case with only theembankment load. Considering that residual settlement isoften discussed on the order of around 10 cm, the overlay
factor can be regarded as an important factor in theexamination of residual settlement.
Analysis was conducted for other two cases using theabove method, and the schedule and number of overlay
repairs for respective cases were found (Figure 7). In thecase where residual settlement was 30 cm three years after
placement in service, six repairs were necessary in 10 years.
In the first four years, in particular, repairs had to be

222

Figure 6. The effect of additional overlay pavement from
repairs in increasing embankment load and settlement

number of overlay pavement(times)5. RESULTS OF THE MAIN STUDY

7
6

residual settlement 30cm

5residual settlement 20cm
residual settlement 10cm

4
3
2
1
0
1～10

10～20

20～30

30～40

40～50

service period of time (years)

Figure 7. The number of overlay pavement repairs forrespective cases6. CONCLUSION
It was found that FE analysis using the Sekiguchi-Ohta model
can express the long-term settlement of peaty soft ground. It
was also revealed that the amount of residual settlement
greatly affects the repair cost.
The authors intend to conduct further studies using themethod applied in this study as an analysis tool and present
the rational amount of residual settlement, repair scheduleand other guidelines.
REFERENCES[1] Civil Engineering Research Institute of Hokkaido,
Manual for Countermeasure against Peat Soft Ground,
pp.42-44, 2002. (in Japanese)[2] Sekiguti, H. and Ohta, H., Induced anisotropy and timedependency in clays. Proceedings of the 9th ICSMFE,
pp.229-239, 1977.
[3] Nishimoto S. and Hayashi, H., A Case Study on The
Long-term Settlement and Life Cycle Cost of A Road on
Peaty Ground, Proceedings of 1st International
Conference on Transportation Geotechnics, pp.397-400,
2008.
[4] Hayashi, H., Mitachi, T. and Nishimoto S., Permeability
Characteristics of Peaty Ground, Technical reports of
Hokkaido branch, JGS, Vol. 46, pp.31-36, 2007a. (inJapanese)[5] Noto, S., Revised Formula for Prediction of Settlement on
Soft Peaty Deposits. Monthly Report of Civil
Engineering Research Institute, No. 446, pp.2-9, 1990.
(in Japanese)[6] Hayashi, H., Mitachi, T. and Nishimoto S, Determination
Procedure of Soil Parameters for Elasto-plastic FE
Analysis of Peat Ground, Proceedings of the 13th AsianRegional Conference on SMFE, pp.145-148, 2007b.

223

50

overlay repair cost
(million JPYen)

conducted every year. In the cases where residual
settlement was 10 and 20 cm, however, the number of
repairs was half or less. In the case of a two-lane road,closure is necessary even for a simple repair like overlaying.
It is considered difficult to gain the understanding of users
and related organizations if repairs requiring road closuremust be made frequently soon after the opening of the road.
On the Mihara Bypass (1.4 km long), box culverts areplaced at average intervals of 170 m. On the assumptionthat residual settlement would cause bumps before and after
these culverts, the relationship between residual settlement
and the overlay repair cost for the length of 1.4 km in 50
years was estimated (Figure 8). The conditions for theestimation were set using the actual repair cost found by
Nishimoto and Hayashi [3] as a reference. Compared withthe case of 30 cm residual settlement, the repair cost wasapproximately 72 and 59% in the cases of 20 and 10 cm
settlement, respectively. It can be seen that the setting of theamount of residual settlement greatly affects the repair cost.

40
30
20
10
0
30cm

20cm

10cm

residual settlement

Figure 8. The relationship between residual settlement andthe overlay repair cost[7] Iizuka, A. and Ohta, H., A determination procedure of
input parameters in elasto-viscoplastic finite element
analysis, Soils and Foundations, 27(3), pp.71-87,1987.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Seismic behavior of pile group in soil slopes: 1g shake table tests
H. Elahi1, M. Moradi2,A. Ghalandarzadeh 3 and A. Elahi 4
SPT Co, Tehran, Iran, 2,3Tehran University, Tehran, Iran

1,4

ABSTRACTA number of 16 physical model tests were conducted using a 1g shake table in order to investigate the seismic behaviorof pile group in a soil slope. The instrumented physical models were exposed to sinusoidal motions of constant
amplitudes. The parameters of the amplitude of the incoming motion, the number and spacing of piles in the group werealtered. Single pile and group piles of 2×1 and 2×2 configuration with pile spacing to diameter ratios of 3, 5.4 and 7.8
were considered. Some observation and conceptual results about piles and pile cap displacements are presented as wellas lateral soil pressure on rear and front piles in group. The most essential result of these set of tests was to support theconcept of utilizing the Newmark sliding block theory in the analysis of group pile within a slope. As the results ofthese tests, the preliminary diagrams for estimation of yield acceleration of pile group in the slope and its reliance upongroup parameters are presented.
Keywords: pile group, seismic, slope, shake table, Newmark methodmeasuring and control system. The model box isconstructedwith Plexiglas which has dimension of 180*80
1. Introduction
2
For many years seismic behavior of pile group is attended cm in plan and 120 cm height (Fig. 1).
from researchers around the world. Against of wide Many sensors such as accelerometers, electrical strain
investigations, complexity of their behavior in soil gauges,displacement gauges (LVDT) are used formakesmany ambiguities in this field of foundation instrumentation which all of them have dynamic samplingengineering. In earthquake loading andpresence of ability. Two dynamic data logger instruments with 20
nonlinear soil-pile-structure interaction, make this issue channels are applied to gathering required data related
more sensible. Furthermore when the piles located in slope with piles, pile cap and soil parameters (Fig. 1).
the complex behavior of these geotechnical structures is 2.2. Dimensional analysis and scaling laws
increased. Often pile group seat on soil slope such as
Reference [2] presented the non-dimensional parameters
foundation of structures in slopes, foundation and/or
for modeling of lateral pile behavior in shake table asabutment of bridges pile supported wharves and somefollowing equation:
roads or pipelines on slopes. According to L p ρ p E p I p ω 2 u 0 mc yabovementioned complexities, there is not any simple and (1)= K  , ,,,4reliableanalytical method in these conditions.Duedg ρ s u 03  u 0 ρ s Gs u 0
tounknown behavior, numerical methods are not reliable
too and their resultsmay not be explained suitable. In which, y: Lateraldynamicdeflection of pile, d: pile
Therefore attempting on exploring seismic behavior of pile diameter, L : pile length, u : displacement amplitude ofp0group in soilslope can be an important and applicable issue input motion, ρ : pile density, ρ : soil density, EpIp: pileps
for investigation. Present work is done as an initiative
flexural stiffness, Gs: shear modulus of soil, ω: frequency
research in this field beneath performing some shake tableof input motion, m0 : mass of pile cap. Piles spacingto
model tests.
diameter ratio (s/d) is an additional non-dimension
parameter which considered pile group effects. Therefore
2. Shake table model tests
the scale factors for pile group shake table models can bepresented as table 1.
2.1. Shake table specifications
Used equipment has several parts such as a table withdimension 180*120 cm2, hydraulic actuator, power unit,

LVDTs

Strain guage
Fig. 1.Schematics of physical models and testing instrument

225

Table 1.Scale factors for mechanical parameter in 1g
model of pile group
Parameter
Lp
EpIp
mc
a0
F
s/d
6.5
4.6
0.75
Scale Factor
n
n
n
1
1/n
1

amplitude and frequency content of the motion [2]. RMS
acceleration is defined as following:

P

P

P

2.3.Geometry of model
Geometry of models is shown in Fig. 2(a). All models areconstructed with 9 layers and total height of 90 cm. Piles
length are 90 cm and the height of slope foundation is 40
cm.
0.18
0.4
0.3

Acceleration (g)0.16

Amplitude (g)

0.14
0.12
0.1

0.2
0.1
0
-0.1
-0.2
-0.3
-0.4

0.08

0

1

2

3

4

0.06

5

6

7

8

9

10

t (s)

0.04
0.02

b

0

0

5

10

15

20

25

30

35

40

Frequency (Hz)
D3

a

D4

acc4

D5
2.1

SG1.1

2.2

SG1.2

2.3

acc3
SG1.3

2.4

SG1.4

2.5

SG1.5

2.6

SG1.6

1 Td 2a (t ).dt
Td ∫0

a RMS =

R

(2)
In which, T d : input motion duration, a(t): acceleration.
Because of less dependency of a RMS to large accelerations
and frequencies and existence of motion duration in it, it isa good parameters for research and engineering
applications [3]. a RMS is used in the current study as
representative acceleration parameter.
2.6. Tests plan
Some of model parameters such as number and arrange ofpiles in group, piles spacing to diameter ratio, slope angle
and input motion specifications such as maximum
acceleration are changed to study seismic behavior of pilegroup in slope. Plan of tests is illustrated in table 3.
2.7. Construction procedure of the models
Construction of the physical model has four importantparts; making a pile group, making a slope with sand rain,
installing the piles in the slope and completing the
instrumentation. Fig. 3 shows above mentioned sequences.
3. Tests results
Piles bending moment, displacement and acceleration ofpile capare recorded and analyzed. A sample of recorded
data is presented in fig. 4.
3.1. Pile cap behavior
Seismic response of pile cap and acceleration of pile cap
are imparted as basic parameters of pile cap response.
A. Pile cap acceleration: Based on try and errorwith
severalexaminations
onvariousparameters,
R

R

R

R

R

R

a RMS.cap / n.s.f 2 was selected in which a cap is pile cap

acc1

Fig. 2.a) Geometry of physical models(dimensions are incm), b) a sample of time history of input motion with
Fourier spectrum
2.4. Materials
a. sandy soil: Firouzkouh (No. 161) siliceous sand is used
for model tests. This kind of sand has uniform aggregationand in the current study it has relative density of about
60%. Some specification of used soil is presented in table2.b- Pile andpile cap: model piles made from Aluminum
pipe with 15 mm external diameterand 1 mm thickness.
Piles tip are conic with 30 degree angle. In each 2*2 group
2 piles are instrumented with 6 strain gauges. Pile cap ismade from steel plate which contains 3 sets of 4 number of
16 mm diameters holes that covered pile spacing todiameter ratios included 3, 5.4 and 7.8.
2.5. Input motions
Base input motions are harmonic with fixed amplitude andfrequency of 10 Hz. Each of physical models was tested
with 2 differentmotions until reach to largedeformations.
At first a weaker motion with maximum acceleration of
about 0.15g exerted to model and then a stronger input
motion with maximum acceleration of about 0.3g wereapplied(Fig. 2(b)). There is a usefuldynamic parameterthat used in previous researches for considering both

acceleration, n is number of piles in group, s is piles
spacing in group and f isinput motion frequency. Fig. 5(a)shows relation between these parameters and base input
acceleration.
B. Lateral displacement of pile cap: Residual and cyclic
displacement are two component of cap displacement at
each time (Fig. 5(b)). To understanding some basic
components of behavior two hybrid parameter including
Disp (Re s ) sn.Disp(Res)/s
andversus(n.
) /( )
Disp (Cyc) d

a RMS .base / sf 2 is shown in Fig. 6.
Shape of cap residual displacement in Fig. 5 (b) shows it
may be possible to estimate cap residual displacement
using Newmark sliding rigid block analysis [4]. Try anderror analyses were conducted to estimate yieldacceleration (a y ) for each test via comparison between
measured and calculated displacement of pile cap (i.e. Fig.
7 for tests 2-3-2 and 1-1-2). Relationship between
calculated a y and some parameters such as baseacceleration (a base ), cap acceleration (a cap ), cap inertial
force (ma cap ), number of piles in group (n) and s/d wasevaluated through try and error progress with varios nondimensional hybrid parameters. Results show that a y isdependent to s/d, a base and a cap (Fig. 8). As seen, a y and(ma RMS.cap ) /(n.s / d ) parameter have the best correlation
R

R

R

R

R

R

R

R

R

R

R

R

(Fig. 9).

226

R

R

R

R

R

R

Table 2. Specification of FIRUZKUH 161 sand
USCS
Name
SP

D10

D30

D50

D60

Passing
#200

Sand

(mm)

(mm)

(mm)

(mm)

%

%

0.16

0.21

0.27

0.3

1

99

Cu

Cc

φ

emax

emin

0.874

0.548

degree

1.87

0.88

400

Table 3.Parameters and specifications of physical models
No.

Test Name

amax (g)aRMS (g)

Slope

s/d

1
2
3
4
5
6
7
8
9
10
11
12
13
14
15
16

1-1-1
1-1-2
2-1-1
2-1-2
2-2-1
2-2-2
2-3-1
2-3-2
3-1-1
3-1-2
3-2-1
3-2-2
3-3-1
3-3-2
3-4-1
3-4-2

0.22
0.48
0.15
0.46
0.22
0.46
0.11
0.39
0.16
0.44
0.20
0.43
0.18
0.43
0.17
0.50

0.08
0.22
0.04
0.20
0.08
0.23
0.09
0.20
0.07
0.21
0.08
0.21
0.06
0.16
0.06
0.23

1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
1:2
horizontal
horizontal

0
0
5.4
5.4
7.8
7.8
3
3
3
3
5.4
5.4
7.8
7.8
7.8
7.8

Group
Configuration
singlesingle
2*1
2*1
2*1
2*1
2*1
2*1
2*2
2*2
2*2
2*2
2*2
2*2
2*2
2*2

Acc4

1.00
0.50
0.00
-0.50
-1.00
0

1

2

3

4

4

5

6

7

8

9

10

SG2.4

2
0
-2
-4
-6

2

Cap Horizontal Displacement (cm)Moment (kg.cm)

Acceleration (g)Fig. 3. Construction of Models

1.5

1

0.5

0

-0.5

-8
0

2

4

6

8

10

0

2

4

Time (s)

6

8

10

12

Time (s)Fig. 4.Sample of recordedacceleration, pile bending moment and horizontal displacement of pile cap (test 2-3-2)
12

D5
Res.D5
Cyc.D50.4
y = 1.46x - 0.03
2
R = 0.88

8

0.3

6

aRMS(cap)/s/f2/n

Cap Hor. Displacement (mm)

10

4
2
0

0.2

0.1

-2

(b)

(a)

-4

0

200

400

600

800

1000

1200

Time (0.01s)

0
0.0

0.1

0.2

0.3

aRMS(base) (g)Fig. 5. a) Cap versus base acceleration, b) components of cap displacement in test no. 3-1-2

227

0.4

n.Disp(Res)/s

5Linear (n.Disp(Res)/s)4

y = 8.47x - 0.07
R2 = 0.87

0.7

n.Disp(Res)/sn.(Res/Cyc)/(s/d)

0.8

3
2

Slope
Hor. Ground1

Linear (Slope)2

R = 0.89

0.6
0.5
0.4
0.3
0.2
0.1

0

0
0.0

0.1

0.2

0.3

0.4

(b)

aRMS.base/(sf2)

0.5

0.0

0.1

0.2

0.3

0.4

0.5

(a)

a RMS.base/(sf2)Fig. 6. a) Variation of n.Disp(Res)/s versus base acceleration, b) variation of cap displacement components versus baseacceleration
2.5

1.8
1.6

2

1.4
1

disp (cm)disp (cm)1.2
calculated

0.8experimental

0.6

1.5
calculated

1experimental

0.5

0.4
0.2

0

0
-0.2

-0.5
0

2

4

6

8

10

12

time (s)

14

16

(b)

18

20

0

2

4

6

8

10

12

14

(a)

time (s)

16

18

20

Fig. 7. Back calculating of yield acceleration (a y ) for a) test 1-1-2, b) test 2-3-2
R-

3.2. Pile Behavior
Pile deflection, moment and shear as well as lateral soilpressure on piles are known as pile seismic response.
According to importance of lateral soil pressure on front
and rear piles (Fig. 10a) and its unknown aspects, test
results on this parameter are presented in the current paper.
Discussion on other mentioned parameters is presented in
4. Conclusion
16 1g shake table model tests on 2*1 and 2*2 pile group in
soil slopewereperformed and the following results about
seismic behavior of pile group in soil slopeare seen andpresented:
Pile cap motion in slope has 2 components: residual andcyclic.
With increasing the base acceleration, a RMS.cap /(n.s.f 2 )which is composed of pile cap acceleration ( a RMS.cap ),-numbers (n) and piles spacing (s) and input motion
frequency (f), increased.
With raising the base acceleration hybridparameter(
a RMS.base / sf 2 ),another hybrid parameters of horizontal
motion of pile cap [ n. Disp(Re s) / s andDisp(Cyc) dandDisp(Cyc)/s] are increased.

R[5]. In the current study maximum lateral soil pressure isnormalized by dividing toz and getpressure coefficient (K). Comparison of lateral soilpressure coefficient of rear and front piles is shown in fig.
10b in which variation of K rear /K front is presented versus ahybrid parameter related with base acceleration and groupgeometry (a RMS(base) .(s/d)).
Soil pressure on rear pile, is 1 to 2 timesof pressure on
front pile. The ratio of coefficients of lateral pressure on
rear and front pile (Krear/Kfront) is decreased whena
hybrid parameter based oninput motion acceleration andgroup geometry [a RMS(base) .(s/d)]is increased.
Results of present tests, confirm the applicability of
Newmarksliding block method in simulation of grouped
piles located in slopes.
Yield acceleration for application of Newmark method forpile group in slopes is different with common yieldacceleration in slopes. Meanwhile, yield acceleration ofpile group in slope (a y ), is increased with raising the
hybrid variable ( ma RMS.cap ) /( n.s / d ) included pile cap
R

R

R

R

R

R

R

R

R

R

inertial force, number of pile in group and piles spacing todiameter ratio.

n.Disp(Res)/s

228

0.25

0.4

0.3
2x1

2x1

2x2
ay slope

0.2

2x2

SingleaRMS.base~0.2g

0.3

Single0.2

ay slope

ay slope

0.2

aRMS.base~0.1g
0.1
2x1

ay1 (g)ay1 (g)ay1 (g)

0.15

0.1

2x2
0.05

0.1

Single

0

0

0

0

10

20
n.s/d

30

40

0.0

0.1

0.2

0.0

0.3

0.2

0.4

0.6

0.8

1.0

aRMS.cap (g)aRMS.base (g)Fig. 8. Relationship between a y and base and cap acceleration, n.s/d in model tests
R

R

0.25

0.25

0.2

0.2

2x1
0.15

0.15

ay1 (g)ay1 (g)

2x2
Single

0.1

0.1

0.05

0.05

R2 = 0.75
0
0.00

0.05

0.10
0.15
ma RMS.cap /(n.s/d)

0
0.00

0.20

0.05

0.10
ma RMS.cap /(n.s/d)

0.15

0.20

Fig. 9. Dependency of a y to group parameters
R

R

2.5

2x1
2x2
Power (2x1)

Krear / Kfront

2.0

Power (2x2)

1.5
R2 = 0.80
1.0
R2 = 0.73
0.5(b)Front Pile

0.0

Rear Pile

0.0

0.5

1.0

1.5

2.0

aRMS(base).(s/d)(a)
Fig. 10. a) Rear and front pile definition, b) relationship between the ratio of lateral soil pressure coefficient of rear andfront piles and a hybrid parameter includes base acceleration and s/dReferences[1] Gohl, W., (1991), “Response of Pile Foundations to
Simulated Earthquake Loading”, Ph.D Thesis;[2] Sabermahani, M., (2008), "A study on seismic
displacement of reinforced soil walls",Ph.D thesis, Tehran
university, Tehran, Iran;[3] Kramer, S. L., (1996), “Geotechnical Earthquake
Engineering”, Prentice Hall, USA.
[4] Newmark, N.M., (1965), “Effects of Earthquakes on
Dams and Embankments”, Geotechnique, 15 (2), pp. 139159.
[5] Elahi, h., (2009), "Seismic behavior of pile group in
soil slope",Ph.D thesis, Tehran university, Tehran, Iran.

229

Effect of Acid Rain on Soil Consistency
Zeinab Bakhshipour ⁽¹⁾, Bujang Kim Huat⁽1⁾, Afshin Asadi⁽1⁾, Shaharin Ibrahim (2), Majid Mirzaei (1)Civil Engineering Department, Faculty of Engineering, University Putra Malaysia,
43400 Serdang, Selangor, Malaysia (1)Environmental Sciences Department, Faculty of Environmental, University Putra Malaysia,
43400 Serdang, Selangor, Malaysia (2)ABSTRACT
Acidic deposit that fall to Earth from the atmosphere is named acid rain. Acidic rainfall as an environmental factor
has a potential to effect of consistency parameter of soil. This paper describes the degree of difference in theproperties of consistency soils in Malaysia of acid rain equivalent to 0.005N of H 2 SO 4 or HNO 3 . The sampling ofsoil from two locations that are Batu Cave (BC) as clay soil and Taman Ukay Perdana (TUP) as weathering granite
soil having different plasticity were composed and distinctly preserved with acid sulfuric (H 2 SO 4 ) and acid nitric(HNO 3 ) to examine the outcome of acid rain on their texture. The amounts of deviation in the engineering properties
are experiential to be more distinct in soils with low plasticity than those with high plasticity.
Keywords: Acid rain, Soil, Consistency, H 2 SO 4 , HNO 3 .
acid rain on soil consistency with different pH 3.6,
4.2, 5.2 and 6.7 in tow kind of soil. Two kind of soil
Clay and granite having different plasticity werecollected and separately acidified with acid rainequivalent to 0.005N of H 2 SO 4 and HNO 3 . In thispaper some of the results obtained are discussed.

1. INTRODUCTION
Environmental Geotechnology is developing as aninterdisciplinary discipline, directing at predicting,analyzing and solving the geotechnical problems
linking the influence of environmental factors [1].
With rapid and developed industrialization of world,
so have become an increase acid rain and an
environmental concern. Mutilation to the naturalenvironment, mainly the bio-sphere, such asalterations in the chemistry of surface water,
reduction in the population of aquatic fauna and flora,
and a growth in the outflow of heavy metals inindustrial country have been blamed on acid rain [2],[3], [4] and [5]. Malaysia is also not safe acid rain
phenomenon. Anthropogenic emissions haveincreased over the year as the level of pollutant
suspended in the atmosphere is increasing.

2. METHODOLOGY
The soil samples used in this study were obtained
from Batu Cave (BC) as clay soil and Taman Ukay
Perdana (TUP). All soil samples were oven dried andsieved BS. The sample of sieve material was used for
additional experimental work. The consistencies ofthe samples were determined for the soil fraction
passing 425 μm sieve. The results are summarized inTable 1.
Table 1. Characteristics of original soil samples ofclay (BC) and granite (TUP)

Acidification of rain is produced by air pollutants
comparable sulfur dioxide and oxides of nitrogen,which decreases its pH as low as 5.2. The process ofeffect of acid rain on soil properties is the H+ content
of acid rain falling for a short duration is very low
compared to cation argument capacity of soil. This
will definitely increase the rate of leaching of cations
from the soil. Fascination of H+, SO42-, NO3- andCO32- by the soil particle is effect on change
physical, chemical and engineering properties [6].

characteristic

One of the greatest important environmental factors,
directly affect of acid rain on soil properties.
Objective of this study was to find degree variation of230

Soil sample
BC
TUP

Liquid limit (W L ) %

52.56

46.50

Plastic limit (W p ) %

55.57

28.58

Shrinkage limit %

10.71

2.14

45

1 kg of each soil sample BC and UTP was separately
shaken for 8 hours with 1 liter of 0.005N H 2 SO 4 and
HNO 3 . The H 2 SO 4 treated samples (BCS, UTPS)
and the HNO 3 treated samples (BCN and UTPN)
were then filtered and air dried. These samples werethen used to find their consistency as per British
Standard [7].

40
Plastic limit (Wp) %2.1 EXPERIMENTAL METHODS

30
25

4

BCS

12

4.5

5
pH

5.5

6

6.5

7

BCN

TUPS

TUPN

10
Shrinkage limit (SL) %

TUPS

3.5

In figure 3 shows the degree of variation of shrinkage
limit (SL) in BCS, BCN, TUPS and TUPN withmaximum LS in pH 3.6 (H 2 SO 4 ) with value of
10.28% and the least value in pH 6.7 is 1.07%.

The degree of variation with liquid limit of BCS,
BCN, TUPS and TUPN was found to be maximum
W L in pH 6.7 (H 2 SO 4 ) with value of 64.61% while
the least in pH 3.6 with value of 46.79%. The degreeof variation in pH 3.6 (H 2 SO 4 ) was shown with valueof 31.24% (Figure 1).

TUPN

8
6
4
2
0

60
Liquid limit (WL) %TUPN

Figure 2. Characterization of plastic limit in clay andgranite in different pH with H 2 SO 4 and HNO 3

Values of Liquid limit (W L ), Plastic limit (W P ) and
Shrinkage limit for the soil samples with different pH
before and after acid treatment are presented in
Figure 1, 2 and 3. The results show that on treatment
with the acids the value of W L , W P and SL got
reduced while that of pH decreased.

BCN

TUPS

35

3

3.1 CONSISTENCY CHARACTERISTIC

BCS

BCN

20

3. RESULT AND DISCUSSIONS

70

BCS

3

50

3.5

4

4.5

5
pH

5.5

6

6.5

7

Figure 3. Characterization of shrinkage limit in clay
and granite in different pH with H 2 SO 4 and HNO 3

4030

The degree of variation in the consistency
characteristics of the soil depends on factors like typeof soil, electrical charge of exchangeable cation
absorbed by soil particles and concentration of cation
in soil water [6].

20
3

3.5

4

4.5

5
pH

5.5

6

6.5

7

Figure 1. Characterization of liquid limit in clay andgranite in different pH with H 2 SO 4 and HNO 3

4. CONCLUSIONS

The degree of variation with plastic limit of BCS,
BCN, TUPS and TUPN was found to be maximum
W P in pH 6.7 (H 2 SO 4 ) with value of 39.44% while
the least in pH 3.6 with value of 21.29%. The degreeof variation in pH 3.6 (H 2 SO 4 ) was shown with valueof 31.24% (Figure 2).

Acid rain results in deviations in physico-chemical
characteristics of soil. pH of the soil reductions while
the concentration of captivated H+ and SO42- rises.
The grain size of soil distribution will move nearfiner sections when subjected to acid rain. Acid rainaffects on consistency properties of different soils
with different pH. The data found through theexperiments
is useful in evaluating
thecompensations to the soil that could occur due to the231acid rain in future. However the greatness of thedecline will depend on determination and strength ofacid rain. In further research need to study in effectand strength of acid rain on different types of soils isfor an improved evaluation of difference in its
physicochemical, consistency and engineering
properties.
5. ACKNOWLEDGEMENTS
The authors extend their sincere thanks to Professor
Dr. Bujang Kim Huat, Dr. Afshin Asadi, andAssociate professor Dr. Shaharin Ibrahim. Also
extend our sincere gratitude to all the authors whose
publications provided us directional information fromtime to time.
6. REFERENCES[1] Manassero M and Deangeli C, “Education in
Environmental Geotechnics”, Proceedings of Indian
Geotechnical Conference, Vol. 2, Allahabad, India,
2002, pp. 840–848.
[2] Hileman, B., "Acid rain: A rapidly shifting
sence," Environ. Sci. Technol., Vol. 17, No. 9, 1983,
pp. 401-405.
[3] Havas, M., "Red herrings in acid rain research,"
Environ. Sci. Technol., Vol. 18, No. 6, 1984, pp. 176185.
[4] Elzerman, A., "Acidification of southern
Appalachian lakes," Environ. Sci. Technol., Vol. 19,
No. 6, 1985, pp. 552-557.
[5] Henriksen, A., "Sulfate deposition to surface
waters," Environ. Sci. Technol., Vol. 22, No. 1, 1988,
pp.8-14.
[6] Kamon Masashi, Ying Changyun and Katsumi
Takeshi, “Effect of acid rain on physico-chemicaland engineering properties of soils”, Japanese
Geotechnical Society, 37 (4), 1997, PP. 23-32[7] British Standard, 1377-2:1990.

232

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Consideration concerning reasonable modified form of Floating-type
Improved Ground
Hijiri HASHIMOTO1, Satoshi NISHIMOTO1 and Hirochika HAYASHI 1
Civil Engineering Research Institute for Cold Region, PWRI, JAPANABSTRACTIn this study, centrifugal model tests were conducted to clarify the rational improvement width (B) and
improvement depth (H) at which embankment stability as well as a deformation prevention effect for the surrounding
ground can be achieved when rectangular floating-type improved ground is created at the toe of an embankment slope.
The results produced the following findings: the creation of floating-type improved ground (FTIG) helps tostabilize embankments; FTIG can be expected to help prevent deformation of the surrounding ground; and the failure of
improved soil does not have a substantial impact on embankment stability. Even if the B/H ratio of improved soil is thesame, unless it is placed (hereinafter, “embedded”) deeper than the sliding line, it does not contribute to displacement
prevention in the surrounding ground.
Keywords: Four or Five Keywords

1. INTRODUCTION
Most stabilization measures for embankments to beconstructed on soft terrain involve the improvement ofground down to the bottom of the solid bearing layer.
Regarding the form of improved soil, the recommended ratioof the improvement width (B) to the improvement depth (H)(hereinafter, “the B/H ratio”) is set to at least 0.5 to 1.0 toprevent the action of excessive bending stress on the
improved soil. As the B/H ratio increases, a greater
improvement effect can be expected, but there also tends to bea corresponding increase in construction costs.
In view of this, the creation of improved ground that does
not involve treating soil down to the bearing layer(hereinafter, “floating-type improved ground”) is considered
to be more effective. However, the extent to which themethod stabilizes embankments and prevents displacement of
the surrounding ground is unclear. Against this background,centrifugal loading model tests were conducted in this study
to review whether greater B/H ratios in floating-type
improved ground or differences in the improvement type used
help to prevent displacement of the toe of embankment slopes
and the surrounding ground.

in which floating-type improved ground was present at the toe(cases 2 to 5). In these cases, observation was performed toverify whether differences in the B/H ratio (i.e., variations in
the improvement width and/or improvement depth with thesame improved area) would affect the failure conditions of
the improved soil and the deformation conditions of thesurrounding ground. The models had a scale of 1/50 of theactual size, and the tests were conducted in centrifugal
acceleration fields of 10 to 50 G (G: gravitational
acceleration).Paper size 210mm× 297mm of A4 paper.
Table 1 Experimental cases and study items
Case
Improvement
Condition
0
Inprovemen
6
t depth
H（m）
4

1

2
3
4
5
Improvement width B（m）
30
1
2
No Improvement
Case2，3 and 4
Improvement width
B/H=0.17
B/H=0.33

Case3，5Same Improvement Area

B/H=0.5
B/H=0.75※Sizes corresponded to actual scale（50G）※Improvement down to base ground 2(case2,3,4) and down to base ground1 only (Case5)▽: Displacement meter
L1 – L3: vertical
L4: horizontal

2. PURPOSE AND DETAILS OF THE CENTRIFUGALLOADING MODEL TESTS

Embankment

The purpose of the centrifugal loading model tests was toverify whether floating-type improved ground is effective instabilizing embankments and preventing displacement of thesurrounding ground.
Table 1 lists the experimental cases and study items. The
experiments included a case in which no improved soil waspresent at the toe of the embankment slope (case 1) and four

Base ground 1(kaolin)Base ground 2(kaolin)
Sandy layers* Figures in parentheses indicate full-scale values.

Fig. 1 Model ground and measurement positions (case 3)233

Please save to MS Word templates directory. Times New
Roman, Font size 10, normal. Try to avoid Underline or Bold
within texts.
Table 2 Properties of base ground, improved soil andembankment
Item
項目

Value
値

Material
材料

Kaolin clay
カオリンクレイNaturalwater content (%)
自然含水比Wn（％）

62.2（WL×1.5）
3

Densityof soil particles (g/cm
土粒子の密度（g/cm） )3

Fragment(%)
礫 分（％）
基礎地盤
BasegroundGrain
粒 size度

Single-axis
一軸圧縮強さcompression

Improved
改良体
soil

Embankment
盛土

2.738
0

Sand
(%)
砂
分（％）

0

Silt (%)
シルト分（％）

45.1

Clay (%)
粘土分（％）

54.9

Maximum
grain diameter (mm)
最大粒径（mm）

0.075

Base ground 1 (kN/m22)）
基礎地盤1（kN/m

12.5

Base ground 2 (kN/m22)）
基礎地盤2（kN/m(3) Embankment
Embankment materials were made by mixing Toyoura
sand and kaolin at a dry-weight ratio of 8:2 and a watercontent w of 10%. The constituents were stirred using avacuum mixer for 40 minutes, and were then cured for a day
or longer.
The embankment materials were used after being passed
through a 2-mm sieve to give them a uniform consistency.
Each layer was compacted to a finishing thickness of 1 cm
using a small rammer.
2.2 Measurement device
Ground level upheaval, horizontal displacement of theembankment slope toe, and settlement of the embankmentcrown were measured with a laser displacement meters (L1 –L4 in Fig. 1).
Silver line noodles were also installed on the visual surface tosupport determination of the soft ground and embankment
deformation mode.

28.5

Material
材料

Kaolin clay +＋ordinary
Portland cement
カオリンクレイ
普通ポルトランドセメント

Cementcontent (kg/m33）)
セメント配合量（kg/m

220

Water
cement ratio (W/C)
水セメント比（W/C）

1

2
2
Single-axis
compression (kN/m)）
一軸圧縮強さ（kN/m360360（3日養生）(3-day curing)

Material
材料

Toyourastandard
sand:kaolin
豊浦標準砂： カオリン= 8 =: 28:2

Compaction(%)
締固め度（％）

85

3
3
Densityof soil particles (g/cm） )土粒子の密度（g/cm

2.665

3
3
Maximum
dry density (g/cm） )最大乾燥密度（g/cm

1.88

Optimalwater content (%)
最適含水比（％）

10.2

Adhesion(kN/m2）2)
粘着力（kN/m

3.2

Angleせん断抵抗角（°）
of shear resistance (° )31.9

2.3 Centrifugal acceleration
The completed model was mounted on centrifugal
loading equipment, and measuring devices were installed.
Centrifugal acceleration was increased gradually to 10, 20,
30, 40 and then 50 G. The centrifugal acceleration retention
time was two minutes at 10 G, four minutes at 20, 30 and 40
G, and five minutes at the final stage (50 G).
Each period of retention lasted until the deformation of baseground 1 and base ground 2 returned to normal after
centrifugal acceleration was increased to the next step.

2.1

Model groundFigure 1 shows a cross section of the completed model
ground, while Table 2 lists the properties of the base ground,the improved soil and the embankment materials. The methodof construction for each type of ground and the experimental
earth tank installation technique are described below.
(1) Base ground 1 and base ground 2 (clay layer)
The base ground was made by adding water to kaolin clay
(in a dry powder condition) to adjust the water content to 1.5
times the liquid limit, i.e., 62%.
One-dimensional consolidation was then carried out
using an auto-load controller. The final consolidationpressure on base ground 1 was p = 30 kN/m2, and for baseground 2 the corresponding values were p = 60 kN/m2.
The reason for introducing a difference in the pressureloads on base ground 1 and base ground 2 was to provide a
sliding surface at the boundary between the two layers.
(2) Improved ground
Ground was improved by adding ordinary Portland
cement (amount: W = 220 kg/m3) to kaolin clay with a watercontent of 1.5 times the liquid limit and agitating it with amixer.

3 TEST RESULTS AND REVIEW
3.1 Improved ground and ground deformation
Table 3 shows transverse and longitudinal sectional viewsof improved soil at the time when embankment loading wascompleted (i.e., after loading of 50 G). The transverse andlongitudinal cross sections were photographed after
completion of the test (1-G field).
In the deformation of the base ground and embankment
when a 50-G load was applied in case 1, no sliding failure
occurred. However, base ground 1 horizontally deformed,
and the whole of the surrounding ground rose as a result(Table 3 a).
Conversely, the deformation modes of the base groundand embankment for loads of under 50 G in cases 2 to 5 were
small (Table 3 b). In the transverse and longitudinal crosssections, it can be seen that the improved soil curved at theboundary surface between base ground 1 and base ground 2 incase 2 (Table 3 f). Likewise, the longitudinal cross section
shows the occurrence of cracking in the improved soil (Table
3 j). As this cracking was located at the boundary surface
between base ground 1 and base ground 2, it can be inferredthat the sliding soil mass acted on the boundary surface with astrength differential between the base ground layers as the234Table 3 Base ground deformation and improved soil in each case
Case

Embankment
heightTransverse cross section of
improved soil

After 50-G loading

Longitudinal cross section of
improved soil

Rise of the surrounding ground
周辺地盤の隆起a)
水平変形

Horizontal deformationb)

Fractured surface

f)

破壊面

破壊面

Fractured surface

j)

c)

g)

k)

d)

h)

l)

e)

i)

m)

Rise of the surrounding ground
周辺地盤の隆起embankment load increased. This may have caused fracturingof the improved soil.
In the longitudinal cross section of the improved soil incases 3 and 4, none of the cracking seen with the improved
soil in case 2 was found (Tables 3 g, h, k, l). This is because
horizontal soil pressure acted on the improved soil. However,
the improvement was wide, with B/H values of 0.33 and 0.5,
respectively. It is presumed that strong resistance of the
improved soil prevented its failure. There was no cracking in
the improved soil of case 5, and its condition was as sound as
those in cases 2 to 4 (Table 3 m), but the surrounding groundrose (Table 3 e). This was because the improved soil did notreach (i.e., it was not embedded in) base ground 2. The
improved soil may have entered a sliding mode in which thesliding of the soil mass could be prevented.

embankment in cases 2 to 4 were about 30 to 50% smaller
than that in case 1. This indicates that floating-type improved
ground is effective in preventing deformation of thesurrounding ground. However, there was no evident
difference in the displacement prevention effect regardless of
the improvement width (i.e., the B/H ratio). Conversely, thesurface displacement gauge values in case 5 were about thesame as those in case 1. This was because little sliding soil
mass prevention effect was achieved without improved soil
embedded in base ground 2.
Next, cases 3 and 5 with the same improved cross section
were compared. The results suggested that the surface
displacement gauge value was reduced in case 3, which
involved improvement deeper than the sliding line (on theboundary surface between base ground 1 and base ground 2).

3.2 Impact on the surrounding ground(1) Surface displacement gauge values
Figure 2 shows the relationship between the embankment
height and surface displacement gauge values, with positive
numbers indicating a rise. The gauge values on the 5-m-high(2) Slope toe horizontal displacement gauge values
Figure 3 shows the relationship between embankment
height and the horizontal position of the embankment slope
toe. Comparison of the horizontal displacement gauge values
in cases 2 to 5 and that in case 1 showed a reduction by about
half in the former. It was thus confirmed that floating-type
235

Accordingly, it was confirmed that floating-type improved
ground would not be able to fully prevent horizontal
displacement of embankment slope toes and/or rising of thesurrounding ground, although it is effective for suchprevention to some extent. Particularly, in case 2 (B/H =0.17), the sliding safety factor (Fs) for circular slippage wascalculated as less than 1 (Fs<1) for an embankment height of
5 m (Figure 4). Although the improved soil fractured in acentrifugal loading model test, the impact on the surrounding
ground was almost the same as in cases 3 and 4, in which the
improved soil was sound. This suggests that failure of
improved soil will not affect embankment stability or
displacement of the surrounding ground. It is also clear that

Surface displacement gauge
value (L1)(mm)
地表面隆起量（L1）
(mm)

400
ケース11
Case
ケース22
Case
ケース33
Case
ケース44
Case
ケース55
Case

300

200

0
1

2

3

4

5

6

盛土高さ heightH p (m) Hp (m)Embankment

1200

法尻水平変位量(L4) (mm)Slope toe horizontal displacementgauge (L4) (mm)Fig. 2 Improvement effect on surface displacement gauge values:
measurement position (L1)
ケース11
Case
ケース22
Case
ケース33
Case
ケース44
Case
ケース55
Case

1000
800
600
400
200
0
0

1

2

3

4

Embankment
heightHp (m)
盛土高さ
H p (m)

5

3.5

ケース1
Case
1
ケース2
Case
2
ケース3
Case
3
ケース4
Case
4
ケース5
Case
5

3
2.5
2
1.5
1
0.5
0
2

3

4

5

6

盛土高さ
(m) (m)Embankment
heightFig. 4

Embankment height and sliding safety factor

4 CONCLUSION
The conclusions drawn from the centrifugal loading
model tests can be summarized as follows:
1) Floating-type improved ground can be used forembankment stabilization.
2) The method also helps to prevent displacement in thesurrounding ground.
3) It was found that improvement soil failure does not
greatly impact embankment stability.
4) Even if the B/H ratio of improved soil is the same, unlessthe improved soil is embedded deeper than the sliding
line, it does not contribute to displacement prevention in
the surrounding ground.
In practical terms, floating-type improved ground is notsuitable for sites where displacement of the surrounding
ground must be avoided. However, if some displacement ispermissible outside the roadway, its application may beappropriate.

100

0even with the same B/H ratio, non-embedded soil will not
contribute to the prevention of such displacement.

Safetyfactor
安全率

ground improvement is effective in reducing slope toe
horizontal displacement gauge values.
The displacement gauge values in cases 2 and 5 wereslightly larger than those in cases 3 and 4. This was because:
1) the improved width in case 2 was smaller (B/H = 0.17),
which reduced the resistance by which soil mass sliding could
be prevented; this in turn caused cracking in the improved
soil, and displacement could not be prevented; 2) in case 5,
soil mass sliding acted on the improved soil, but thisimproved soil was embedded in base ground 2. As a result,soil mass sliding was not prevented in the improved soil.

6

Fig.3 Improvement of slope toe horizontal displacement gauge:
measurement position (L4)5 REFERENCES[1] S. Kuga, T. Hattori, M. Kamimura, N. Sagawa, “Studyon a test embankment on stabilized soft ground,” Japan
Society of Civil Engineers 54th Annual Academic
Conference, pp. 466 – 467, 1999.
[2] K. Tsubota, H. Nakajima, M. Nishigaki, “Study on
measures against settlement of surrounding areas on soft
ground by embankment construction,” papers of theJapan Society of Civil Engineers, Vol. 63, No. 3, pp. 323– 334, 2007.
[3] Public Works Research Center: Design andimplementation of the deep mixing method for work onland, revised edition, 2004.
[4] H. Hashimoto, S. Nishimoto, H. Hayashi, S. Kajitori, “A
rational form of floating-type improved ground,” Japan
Society of Civil Engineers 67th Annual Academic
Conference, 2012.

236

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia,Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Numerical Study Effect of Nail Angle Change on Stability and Displacement of
Excavation Wall in Cohesive and Cemented Non-Cohesive Soils
Navid Shahnazi1, Jahangir Khazaie2, Mohamad Sharifipour3 and Kiarash Ashtari4
Tehran, Iran
ABSTRACT
Increasing population growth and the need for buildings of greater height, and condensation of urban texture,
construction of underpasses in highways intersections with height difference, demolition of old buildings andconstruction of new and high buildings next to each other involve vertical deep excavation. In the meantime, soil
consistency of the excavated walls is considered as one of the crucial issues in geo-technique engineering.
Installation angle of reinforcement member (element) in excavation wall and soil shearing strength effects on nailangle is among the parameters which affects design, installation and execution methods as well as execution costs. Since
parameters of soil shearing strength are amongst the effective elements in obtaining the nail’s optimum angle, and withchange of these parameters, the nail’s optimum angle will change, in this research, it has been tried by numericalmodeling and problem analysis using Plaxis software to design 10-meter excavation with specific parameters ofcohesive and cemented non-cohesive soils, and in the next step, in has been tried to investigate effect of changes insoil’s shearing strength parameter in the 2 states on nail angle as well as on displacement and confidence coefficient of
excavation wall in this type of soils.
.

Keywords: stabilization, vertical slope wall, nail angle, soil shearing resistance parameters
1. INTRODUCTION(w.B.wei, y.m.cheng, 2010) by studying on different
slopes represented that with the increase of nail angle,
first Factor of Safety goes up gradually and then down.

Soil nailing is one of the suitable methods for thestabilization of earth walls which its execution way isfrom up to down contrary to other soil stabilization
methods.

(TohidAhklaghi, Hassan Panahpoor, 2011) in considering
the effective factors in the optimum plan of nailed walls
by using Plaxis Software propounded that in case of using
horizontal nails, changing horizontal shapes in excavation
have less amounts and also the increase of nails angles to
15 degrees causes the safety factor of excavation toincrease.

Reinforcement angle is one of the effective parameters inthe stability of nailed walls. Reinforcement angle (nailing)
plays an important role in improving nail resistance, forthis reason also in practical cases nails should be placedwith deviation toward horizontal axis. In this paper theeffect of changing the various parameters of soil shear
resistance on nail angle will be considered by using
Limited Parts Software to specify how the optimum nailangle varies in granular soils and cohesive ones. In vitro
studies done by (Satyendra Mittal, 2006) showed that
Factor of Safety will decrease with theincrease ofreinforcement angle (nailing) to more than 15 degrees.

2- Analytical ModelIn this paper numerical modeling is done by use of Plaxis
Software. This software is able to do non-linear analysis
and also calculate movement and the safety factor of
excavation wall. For assurance of done modeling primarymodel has been calibrated with the existing model in
Plaxis Manual and then the planned model is done.
Studied vertical slope is 10 meters deep and 15 meters
wide and the effect of load in the movement of excavationwall has been waived. In the contemplated plan, 2 firstnails are 7 and 3 middle nails are 6 and 2 last nails are(Chai- Cheng Fan, Jiun-Hung Luo, 2008) in a numerical
study regarding the optimum arrangement of soil nailingfor the stability of excavation concluded that the optimum
angle of nail decreases with the increase of slope angle sothat zero-degree angle has been introduced as optimum
nail angle for vertical slope.

237

5meters long. Also 6 first excavations are 1.5 meter andlast excavation is 1 meter deep.

3-2- How to Model Nailing
The modeling of the excavation wall is done in Plain
Strain way and 15-node triangular element has been used.
Nails are made of ribbed bar Φ32 and with yield stress
400 Mpa. The excavation diameter of bore is 10 cm andnails are modeled by use of Geogrid element that theshear and flexural capacities of the nails are waived indesign. This performance will be toward assurance. The
below relations are also used for obtaining the equivalentelasticity modulus and axial stiffness of nail.
(Equivalent Elasticity Modulus)
𝑬𝒆𝒒 = 𝑬𝒏 �

Figure 1- The Analytical Model of the Studied Areaand Meshing the
Nailed Wall

(Axial Stiffness)

1-2- Boundary and Meshing Conditions
In this study the dimensions of the analytical model havebeen considered in a way that any location change inborders is zero and just 2 sides of the model will be able
to move in vertical direction. 15-node triangular meshes
are also used for meshing. The general mesh of
atmosphere is Coarse and just nails are refined line twice.

𝑬𝑨 =

EA
EI
d

Average specifications related to the soil of theconsidered location are presented in Table 1 that soil issandy with average cementation. Also Mohr-Coulomb
Standard is used for modeling soils.

PR

P

P

P

P

P

P

Reduction analysis is also used for obtaining Factor of
Safety. In this method the resistance parameters of the soil
heap decrease gradually and finally break happens, in fact
Factor of Resistance Decrease in break time is the same
safety factor of slope.

RP

For considering production stages and its effect on themodel, phasing has been accomplished in 7 steps and ineach stage, according to the execution of the nailed wall,
one-level excavation is done and the nail and shotcrete
cover of that stage becomes active.

Table 1- The Specifications of the Used Soil inAnalysis

The specifications relative to nail are also presented inTable 2. The effects of subterranean water are waived as
nailed walls are generally drained before shotcrete.
Meanwhile nails are modeled in an elastic way.
Nailing Method
Nail Length
EA

4.38× 106 KN/m
1.46× 104 KNm2/m
0.2 m

Table 3- The Specifications Relative to the ShotcreteCover18.6 KN/m 3
18.6 KN/m 3
4 Kpa
30
3×104 KN/m 2
0.28
P

𝑬𝒆𝒒 𝝅𝑫𝟐𝑫𝑯
��
𝑺𝒉
𝟒

Plate element is used for the modeling of the shotcrete
cover. The cover juncture to soil is rigid and modeled inelastic way. The cover specifications are presented inTable 3.

2-2- The Specifications of Materials and Related
Parameters

γ sat
γ dry
c
φ
E
υ

𝑨𝒈
𝑨𝒏
� + 𝑬𝒈 � �𝑨𝑨

3-The Results of Parametric Study on the Analytical
Model

Injection
Variable
1.58×105KN/m
PFirst the effect of the nail angle changes in stability andhorizontal movement of the excavation has been
considered in granular soil.

P

Table 2- The Specifications of the Used Nails inAnalysis

238

Then we make the internal friction angle of soil fixed(φ =30) and change soil cohesion and proceed to consider
Factor of Safety and the horizontal movement of the
excavation in different nail angles in order to get the
optimum reinforcement angle.

movement changes of the excavation wall will decrease
toward the increase of soil cohesion.
2-3- Considering the effect of friction angle on the
optimum reinforcement angle (nail) in soils with low
cohesion

1-3- Considering the effect of cohesion on the optimum
reinforcement angle (nail) in soils with low cohesion

At this stage we consider soil cohesion fixed (C= 4 KPA)
and change the internal friction angle of soil and proceed
to consider Factor of Safety and the horizontal movementof the excavation under different nailing angles.

1.5
1.4

2.000

1.2

1.800

1.1

1.600

FOS

FOS

1.3

1
0.9
0.8

1.200
0

1

2

3

4

5

6

1.000

7

0.800

ΔC , Kpa
Nail Angle 0

Nail Angle 5

Nail Angle 15

Nail Angle 20

28

30

32

34

36

38

40

Δφ , Deg
Nail Angle 0

Nail Angle 5

Nail Angle 15

Nail Angle 20

Nail Angle 10

Figure 4-The Effect of Changing the Internal Friction Angle of Soil on
Nail Angle with regard toFactor of Safety –Soil Cohesion (C=4 KPA)

As regarded in Figure 2, with the slight increase incohesion amount, the safety factor of the excavation wall
will have the greatest amount in return for 20- degree
nailing angle and this process is increasing.

0.05
0.045

0.070

0.04

0.065
0.060

0.035

0.055

Ux

0.050
0.045
0.040

0.03
0.025

0.035

0.02

0.030

0.015

0.025
0.020

26

Nail Angle 10

Figure 2-The Effect of Cohesion Change on NailAngleWith Respect toFactor of Safety

Ux

1.400

0

1

2

3

4

5

6

0.01

7

26

28

ΔC , Kpa
Nail Angle 0

Nail Angle 5

Nail Angle 15

Nail Angle 20

30

32

34

36

38

40

Δφ , Deg
Nail Angle 10

Nail Angle 0

Nail Angle 5

Nail Angle 15

Nail Angle 20

Nail Angle 10

Figure 3-The Effect of Cohesion Change on NailAnglewith regard to the
Horizontal Movementof the Excavation Wall- the Internal FrictionAngleof Soil (φ=30) degree

Figure 5-The Effect of Changing the Internal Friction Angle of Soil on
Nail Angle with regard tothe Horizontal Movementof the Excavation
Wall–Soil Cohesion (C=4 KPA)

As observed in Figure 3, the horizontal movement of the
excavation wall will decrease with the slight increase in
the cohesion amount of soil. Meanwhile it is consideredthat the greatest reduction of the horizontal movement of
the excavation wall occurs in the little nailing angles andwith the increase of the nailing angle, the horizontal

As observed, the optimum nail angle decreases with theincrease of friction angle in soils with a little cohesion sothat the optimum nailing angle is ever zero degree atfriction angle about 35 degrees and more. Furthermore theoptimum nail angle will increase with the reduction of the239

internal friction angle of soil so that the optimum angle isabout 20 degrees.

Modulus, Factor of Poisson, Special Weight of Soil andShotcrete Features and just the shear resistanceparameters (the friction angle of soil and cohesion) havechanged.

3-3- Considering the effect of cohesion on the optimum
reinforcement angle (nail) in cohesive soils

4- Conclusion and Totaling

In this section, the internal friction angle of soil is fixedand equal to (φ= 30) and soil cohesion will change fromC = 15 KPA to C = 25 KPA.

1- The least movement occurs at zero degree incohesive soils but in granular soils with paltry
cohesion the least movement is not at zerodegree angle.
2- Contrary to what is indicated in the papers(Chia-Cheng Fan, Jiun-Hung Luo, 2008)and(TohidAhklaghi, Hassan panahpoor,2011), zerodegree angle has the least horizontal movement,
this matter has no meaning for granular soils
with paltry cohesion.
3- In granular soils and cohesive ones, zero- degree
angle has the least factor of safety and 20-degree
angle has the greatest factor of safety.
4- According to the control standard regarding
restricting the movement of the excavation wall(the maximum horizontal break in amount of
0.005 H) is observed in granular soils that theabove standard is gratified for the studied friction
angle and cohesion interval just at 20-degree
nailing angle.
5- In granular soil, the effect of the friction angle
increase in the increase of the excavation safety
factor is more than the cohesion increase so thatnailing in condensed granular soils can be donein desirably.
6- Based on the obtained results from this research,
15-degree nail angle which is often used bycontractors and consultants in designs, in any of
granular and cohesive soils does not havepreference to other angles.

2
1.9

FOS

1.8
1.7
1.6
1.5
1.4

15

17

19

21

23

25

ΔC , Kpa
Nail Angle 0
Nail Angle 15

Nail Angle 5
Nail Angle 20

Nail Angle 10

Figure 6 -The Effect of Cohesion Change on Nail Angle withregardtoFactor of Safety the Internal Friction Angle of Soil (φ= 30)
degree0.02000.01950.0190Ux

0.01850.01800.01750.01700.01650.0160

15

17

19

21

23

25

ΔC , Kpa
Nail Angle 0

Nail Angle 5

Nail Angle 15

Nail Angle 20

Nail Angle 10

References[1] Byrne RJ, Cotton D, Porterfield J, Wolschlag C,
Ueblacker G. Manual for design and construction
monitoring of soil nail walls. Federal Highway
Administration (FHWA) Report No. FHWA-SA-96-069,
Washington, DC, USA; 1996.

Figure 7-The Effect of Cohesion Change onNail Angle with regardsto
the Horizontal Movement of the Excavation Wall – TheInternalFriction
Angle of Soil (φ= 30) degree

As considered in figure 6, 7, the optimum nail angle
decreases in cohesive soils with considering the horizontal
movement of the excavation wall so that zero degree isthe best amount and the optimum nail angle has the bestamount at 20 degrees with respect to Factor of Safety.

[2] PLAXIS BV. User’s manual of PLAXIS. Publishedby A.A. Balkema Publishers; 2002.
[3] Sivakumar Babu GL, Srinivasa Murthy BR, Srinivas
A. Analysis of construction factors influencing the
behaviour of soil-nailed earth retaining walls. Ground
Improv 2002;6(3):137–43.

At the end it is necessary to remind that the horizontal andvertical distances and lengths of the nails have been
considered fixed in all the above stages. In the meantime
all the soil specifications are constant such as Elasticity240

[4] Fan CC. Effect of geometric layout of soil nails on
stability of soil nailed slopes. In: Proceedings of 15th
Southeast Asia geotechnical conference, Bangkok,
Thailand; 2004. p. 309–14.

[8] Chia-cheng Fan.JIUN-HUNG Luo.Numerical Studyon the optimum layout of soil-nailed slopes. computer andGeotechnics (2008).

[5] S. Mittal, R.P. Gupta and N. Mittal . Housing
Construction On Inclined Cuts Asian Journal Of Civil
Engineering (BUILDING AND HOUSING) VOL. 6, NO.
4 (2005) PAGES 331-346

Numerical Analysis of peformance of soil nail walls
in seismic conditions. ISET Journal of Earthquake
Technology, Paper No. 496, Vol. 45, No. 1-2,
March-June 2008, pp. 31–40[9] G.L. Sivakumar Babu and Vikas Pratap Singh[6] Satyendra mittal. Soil nailing application in erosion
control – an experimental study Geotechnical andGeological Engineering (2006) 24: 675–688[10] Vikas Pratap Singh • G. L. Sivakumar Babu.
2D Numerical Simulations of Soil Nail Walls. Geotech
Geol Eng (2009).

[7] Liew S.S and Khoo C.M., “Performance of Soil Nail
Stabilisation Works for a 14.5m Deep Excavation in
Uncontrolled Fill Ground, Proceeding of 16th Southeast
Asian Geotechnical Conference, Subang Jaya, Malaysia, 8- 11 May 2007, pp. 827-837[11] W.B. Wei, Y.M. Cheng Soil nailed slope bystrength reduction and limit equilibrium methods
Computers and Geotechnics 37 (2010) 602–618

241

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14-16 November 2012

An Investigation On Behavior Of Geogrid Reinforced Soil Retaining Walls Applying
Finite Difference Method (F.D.M)

Jahangir Khazaie1 and Alireza Sadeghabadi2
Assist. Prof. of Civil Eng., Dept. of Civil Eng., Univ. of Razi, Kermanshah, Iran; Tel: +98-9123271926;

1

Fax: +988314283264; Email: j.khazaie@razi.ac.ir
M.Sc. Student of civil Eng., Dept. of Civil Eng., Science & Research Branch., Islamic Azad University,
central Branch(Markazi), Iran;2

Tel: +98-9122070810; Fax: +988314283264; Email: alireza_sadeghi56@yahoo.com
ABSTRACTDue to increasing usage of polymeric reinforced retaining walls in civil projects, researches on the behavior of thesetypes of walls are further essential. So in this paper the behavior of geogrid reinforced retaining walls is surveying by
numerical method (F.D.M). For this purpose FLAC2D (Fast Lagrangian Analysis of Continua) software is applied torepresent numerical model that experimental results are available. Then the software capability is assessed in wall
deformation, reinforcement tension forces and vertical displacement of reinforced soil layers. As a result of thisresearch, way of polymeric materials arrangements is specified according to their strength parameters.
Keywords: Geosynthetic, geogrid, finite difference method “FDM ", Plasticity model, Polymeric reinforced soil
retaining walls, FLAC2D software, Numerical modeling.

1. Introduction
According to worldwide application of retaining walls in many projects, specially urban construction
projects, further study is felt in this case. In this reason, the geosythetic reinforced soil as a polymericcomposite material and an updated reinforcing method has caused a revolution in geotechnical engineering
for its simple usage, cost-saving and ability to improve the visual appearance. Many field and laboratory
studies are performed on different reinforced retaining walls, such as full scale and small-scale model andalso centrifuged decreased models.
Researches in this field- specially bearing capacity of reinforced retaining wall-were proposed in 1960 by“Vidal” for the first time and were being continued till present.

In 1998 “Reda M. Bakeera, Sayed M. Sayeda, Peter Catesa, Rajesh Subramania” studied about
Pullout and shear tests on geogrid reinforced lightweight aggregate. Their researches
Contains the results of laboratory tests conducted on a lightweight aggregate reinforced with a geogrid. The
large-size shear tests included direct simple shear tests on the aggregate, shear tests on the aggregate
reinforced with a uniaxial geogrid, and geogrid pullout tests. All tests were performed on air-dried material
samples. Each type of test was repeated using three different confining pressures representing different
242

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012heights of fill. The angle of internal friction (φ) of the lightweight aggregate material and its interface
friction angle (δ) with the geogrid were found to be 52° and 48°, respectively. These values represent anefficiency of friction angle mobilization (E φ ) of about 87%. Some crushing occurred in the material under ahigher confining pressure specifically near the front edge and center of the shear box.
In 2006, Nicola Moraci, Domenico Gioffrè had a research on,”A simple method to evaluate the pullout
resistance of extruded geogrids embedded in a compacted granular soil”
In 1994 Mr. porbaha performed laboratory study on a “centrifuge modeling of geotextile reinforced
cohesive soil retaining walls” .He had made a .51m height .61m width and 1.12m length model of soil
reinforced soil with 8 geogrid layer embedded on sandy foundation .
And in 2011, by Kerry Rowe & Graeme D, A numerical examination of the behavior of an 8 m highgeosynthetic reinforced soil wall constructed on a layered foundation stratum is described. The foundationconsists of a 0.8 m hard crust, underlain by 2.95 m of soft loam (sandy/silty) and then 1.3 m of stiff clay.
Below the clay is 1.75 m of fine sand underlain by a layer of clayey/fine sand extending to a depth below 10
m. The wall was constructed with 16 segmented concrete facing blocks, a sandy backfill material with 30%
fines and 11 layers of geogrid reinforcement 6 m long. Five additional, one meter long layers of
reinforcement were used between the 6 m long layers within the upper 5 m of the wall to improve the local
stability of the facing blocks. The analysis examines the effect of uncertainty regarding the drained andundrained strength of the loam foundation material, its stiffness, the thickness of this soft layer and its
position with respect to the bottom of the wall on the calculated behavior and compares the calculated andobserved behavior.

2. Model specifications
In a modular block or shotcrete wall, the geosynthetic layers are placed between the concrete blocks
at regular vertical spacing. The blocks or shotcrete are used to retain the backfill. The dead weightof modular blocks contributes to the global stability of the GPSRW. The contribution of the facing
was neglected in early designs; then, Bathurst and Simac (1994) and leshchinsky(1993) proposed
limit equilibrium design to account for the modular blocks.
Here, we consider a geogrid reinforced retaining wall. This model is made in Federal Highway
administration (U.S. Department of transportation-FHWA) in 1996. The wall has a length of 420cm &600cm height. The reinforcements with full length have been laid in horizontal direction in 10 layers of soil.
A linear surcharge pressure q y (=75 KN/m2) acts on the top of the wall by the length of 1.5m. The facing of
this wall is placed in a depth of 6cm into the dense sand soil foundation. The rigid rock dimension is about
33.4×19.5m, as the wall is placed in length of 13.7m inside it. (Figure 1)

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Figure 1) Schematic model of Geogrid reinforced soil retaining wall.

3. Numerical modeling of Full scale Model
The model form is shown in Figure 2. Dimension of 67×34 meshes are variable. Dense sand foundation
modeled as a homogeneous shape in order to minimize eventual errors.

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Figure 2) model of reinforced soil retaining wall in the software
In this episode, the full scale reinforced FHWA model collaborates by applying FLAC2D and experimental
results. The soil is a “Mohr-coulomb” plasticity model.
Modeling follows a construction stages. At first, in this method embankment is done in one layer .and thenpolymer mesh embeds on the soil layer and this period repeats at 8 stages.
It should be noted that the facing panel modeled in two methods, one is applying an elasticity model and theother one is applying a beam structure. Modeling methods are shown in table1 as a summary.
Material

Model

type

soil

Mohr-coloumb------

Facingpanel

Elastic------

Elastic

Beam structure

Elastic

Cable structure

Polymer(geogrid)

Table1) Modeling of different properties in FLAC2D

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4. Material properties
4.1. Reinforced backfill soil properties(1)Young modulus (E): 15000 k.pa
(2)Dry density (γ):

17.6 KN/m2(3) Poisson (ν):0.3(4)Cohesion(C):18.5 KN/m2(5)Internal friction angle(φ):36
4.2. Facing panel properties:(1)Young modulus (E):300GPa(2) equivalent thickness:

17cm(3)Yielding stress (for structural elements): 2400 KN/m2
4.3. “Geogrid” properties:(1)Young modulus (E): 160 M.pa
(2)Ultimate strength tensile: 54 k.pa
4.4. Foundation bed:
(1)Young modulus (E): 25000 k.pa
(2)Dry density (γ):

17.6 KN/m2(3) Poisson (ν):0.3(4)Cohesion(C):19.5 KN/m2

4. Verifying numerical results:
To validate the developed numerical model, as noted in above, a physical model was constructed. The
developed numerical model was used to predict the behavior of the reinforced soil structure in the model
test under applied load. The results of the model performed by this numerical method were compared withthose measured in the experimental test (FHWA-A0).
Diagrams printed in figures 3a, 3b show respectively the horizontal displacement of the facing and verticalsettlement on the top of this wall predicted by the numerical analysis, finite difference analysis, together
246GEOMAT 2012
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with those measured in the experimental model test. Comparison of the results shows that numerical results
predicted by F.D.M are in a good agreement with the experimental results.
As the diagram shows in these figures, when the panel is predicted as a structural element, beam, results arecloser to reality special in deformation of the panel prediction in F.D.M.

Figure 3) Panel deformation in different levels of the wall.

5. A Numerical study in GPSRW
5.1. Displacement:
The maximum settlements occur in the central region (core) of the wall in an specified Hyperbolic
limitation.
Moving away from this area to the regions lead to staddle and upper surfaces, settlement contours decrease.
But in opposite direction to upper regions end up to the external surface of the wall adjacent to loading area,
displacements in Y-direction increases again. (Figure 4-b)

In general, settlements in upper reinforcement layers is bigger than lower ones because failure wedges occur
in a surface with slope angle (45+Φ/2). The maximum amount for displacement contours in X direction
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14-16 November 2012

into the wall occurs in the zones near facing panels and upper layers. Excessive distribution of displacement
contours in the zones near the facing panel and interfaces with foundation is beheld. (Figure 4-b)

Figure 4-b) Settlement contours distribution

Figure4-b) X-displacement contours distribution

5.2. Stresses contours:
Design of a geosynthetic reinforced embankment should basically verify the strain compatibility between
the soil and the reinforcements and ensure that the anticipated displacements are consistent withperformance criteria of the structure. Total horizontal stresses, shows their values in backfills. As shown infigure 5-a, "S xx " in concentrated adjacent to linear loading with minimum amounts. In other zones, contours
distributes as pressure with a moderate incremental trend.
Compressive stresses under linear load indicate that moderate stresses placed within failure wedge, Henceapplication of low compressive strength geogrids is recommended in this region. Also, total verticalstresses occur in a parabolic limitation, consist of the soil wall and the foundation precisely placed under it.
As shown in figure 5-b, maximum compressive stresses start from a central region and contours spread with
reduction proceedings. It's noted that there're maximum tensile stresses adjacent to linear loading region,
again.
Additionally, the basic formulation for Finite Difference Method" (FLAC) is for a two dimensional
plane-strain model. This condition is associated with long structures or excavations with constant crosssection and acted on by loads in the plane of the cross section.

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Figure 5-a) stress distribution within geosynthetic reinforced soil retaining wall (σx , F.D.M)

Figure 5-b) stress distribution within geosynthetic reinforced soil retaining wall (σy, F.D.M)
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Construction materials and Environment.
14-16 November 2012

5.3. Displacement of reinforcements:
Diagrams printed in Figure 7 show vertical displacement of reinforcements in "mm". These displacement
curves are decreasing from the upper layers to lower ones.
Figure 6) vertical displacement ofreinforcements as
Difference Method

resultofFinite5.4. Forces in reinforcements:
Numerical results Finite Difference Analysis for tension forces in polymeric reinforcements was being
printed in a diagram. As shown in this diagram, tension in these types of geogrids increases from base of thesoil wall to the surface. As regards, tension is further adjacent to elastic panel (figure 8).

Figure 7) Tension in reinforcements
diagrams by distance from panel as
result of FLAC2D.

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14-16 November 2012

5.5. Reinforced soil elasticity modals:
Normalized maximum axial forces in Geogrids for different values of Elasticity modulus by the different
height of the reinforced wall are shown in figure 8.
As shown in this figure the values of normalized maximum axial forces reduce with increasing Elasticity
modulus. Also, with increasing Elasticity values modulus from 10 M.Pa. to 70 M.Pa normalized axial
tensions reduced to 30%.However,with increasing values of “E” from 70 M.Pa. to upper values normalized
axial tensions decrease with a reduction rate.

Figure 8) Normalized maximum axial tensions by different height of the wall with variable
Elasticity modulus.

6. Conclusion
1-Finite Difference analysis were conducted to simulate the construction response of a geosynthetic
reinforced soil retaining wall with a concrete block or shotcrete facing.
2-Comparison between measured and predicted behavior are presented for the wall deformation, verticaland lateral stresses in geogrid layers.
3- This study presented a Finite Difference analytical model, based on the theory of plasticity which can beused in analysis of geogrid reinforced soil structures.
4- Axial Tension forces in these type of geogrids increase from base of soil wall to the surface, so the upper
layer reinforced layers sustain greater tensions.
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14-16 November 2012

5- Polymeric materials manufactured by "Tenser" company are convenient to reinforced soil walls.
6- Linear loading along the wall surface affects larger settlements in the layers near the surface.
7-"FLAC2D" software (based on F.D.M) is verified by collaborating with experimental results.
8- The values of normalized maximum axial forces reduce with increasing Elasticity modulus of thereinforced system.

References
1. FHWA (2001), mechanically stabilized Earth walls and reinforced soil slopes. Design and construction
Guidelines, FHWA-NHI-00-043, Federal Highway Administration, U.S. Department of Transportation.
Washington D.C.
2. Rowe, R.K. and skinner, G.D .(2001) Numerical analysis of geosynthetic reinforced retaining wallconstructed on a layered soil foundation, Geotextile and Geo-membranes, 19(7), 387-412.
3. FLAC's Manual, Itasca Consulting Group, Inc. Minneapolis, Minnesota (version 4.0)
4. FHWA,(1998)”Geosynthetic Design and construction guideline” publication No, FHWA-HI-95038,National Highway Institute.
5. Bathurst, RJ.,Walters, D.,vlachopoulos N.,burgess, P., and Allen T.M.,(2000),”Full scale Testing of
Geosynthetic reinforced walls”, ASCE special publication, proceeding of GeoDenver, PP.201-217
6. Bowels, Joseph E. Foundation Analysis and Design-5th ed. McGraw-Hill, New York,1996.

252

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effects of humic acid and salt additives on the behaviour of lime-stabilisedorganic clay.
1.

N.Z. Mohd Yunus, Department of Geotechnics and Transportation, Universiti Teknologi Malaysia, Skudai.
2. D. Wanatowski, and L.R. Stace, Faculty of Engineering, University of Nottingham, United Kingdom.

ABSTRACTThe presence of organic matter, in particular humic acid, is one of the most important factors affecting the effectivenessof lime stabilisation of clays. In this paper, an investigation on the effects of humic acid content on the strength oflime-stabilised organic clay is presented. Artificial organic clay with different humic acid contents of 0.5%, 1.5% and3.0% was used. The organic clay was stabilised with 5%, 10% and 15% of hydrated lime. The strength of thelime-treated organic clay was analysed using the Unconfined Compression Strength (UCS) tests at different curing
periods of 7, 28 and 90 days. A microstructure analysis of selected specimens was also carried out using a ScanningElectron Microscope (SEM). The results showed that a significant strength loss was obtained beyond 1.5% humic acidcontent and the strength was diminished at longer curing periods. However, it was found that the behaviour oflime-treated specimen with 1.5% humic acid content can be successfully improved by addition of 0.5% chloride salts.
Furthermore, the shear strength of the specimens increased with increasing curing period. Therefore, there is a realpotential for the use of chloride salts to improve the lime stabilisation process of organic clay in geotechnical practice.
Keywords: organic clay, lime stabilisation, salt additives, shear strength, laboratory tests

1.

INTRODUCTION

2.

Lime as a chemical stabiliser has been proved successful withregards to its ability to alter the properties of weak soils,
especially those of a cohesive nature [1], [2], [3] and [4]. In
contrast, lime stabilisation of organic clay has always faced acomplex outcome in the long term due to the presence of highconcentrations of organic matter in clay soil [5], [6], [7] and[8]. Specifically, among the constituents of organic matters,
humic acid is believed to be the most disruptive to the lime
stabilisation process. In fact, it has been reported that morethan 1% humic acid content in clay may render the lime
stabilisation process ineffective [9], [10], and [11]. Therefore
for organic soils, lime alone may not be effective or evensufficient as a stabiliser.
One way of mitigating the adverse effects of humic acid isby introducing a small amount of chloride salts (i.e. CaCl2
and NaCl) to lime-treated organic clay. The idea of adding
admixture to lime-treated clay is to provide additional cations(i.e. Ca2+ and Na+) to promote ion exchange activity anddissolution of clay minerals such as silica and alumina for
pozzolanic reactions. These reactions are thought to enable
the greater development in shear strength of organic clay,
especially in the long term [9], [12], [13].
This paper presents an experimental study on the behaviourof lime-treated organic clay with various humic acid contents.
In addition, the influence of 0.5% CaCl2 and 0.5% NaCl on
the behavior of lime-treated organic clay is investigated.
Changes in the soil structure due to the addition of chloride
salts were also examined using the Scanning Electron
Microscope (SEM).

MATERIALS

Artificial organic clay used in this study was prepared bymixing commercial kaolin with commercial humic acidcontents of 0%, 0.5%, 1.5% and 3.0% according to the dry
mass of kaolin. The chemical elements present in each soil
mixture are given in Table 1. In general, the effect of humic
acid on the chemical elements can be detected from theincrease in carbon and oxide ions. As a result, the increase inamounts of carbon and oxide ions and the decrease in silica
and alumina contents can be observed with increasing humic
acid content, as shown in Table 1.
Table 1. Chemical elements in the tested soils.

253

Weight %Element

0.5% humic acid

1.5% humic acid

3% humic acid

C

2.8

2.74

20.16

O

50.45

54.54

48.97

Na

0.36

0.41

0.81

Mg

0.65

0.64

0.56

Al

16.94

18.47

12.8

Si

24.74

21.52

14.07

P

0.18

0.28

0.16

S

0.05

0.12

0.42

Cl

0

0.01

0.19

K

3.42

0.79

0.59

Ca

0.02

0.04

0.33

Ti

0

0.02

0.18

V

0

0.01

0.01

Fe

0.39

0.4

0.74

The results of index testing carried out on inorganic andorganic clays are summarized in Table 2. All of the tested
soils were suitable to be stabilized by lime as the plasticity
index (PI) is more than 10.
Table 2. Physical properties of clay with different humic acidcontents.
Property
Liquid limit (%)
Plastic limit (%)
Plasticity index (%)
Specific gravity
pH
OMC (%)MDD (kg/m3)

3.

Humic acid content (%)
0.5
1.5
64.4
63.6
33.0
33.8
31.4
29.5
2.53
2.51
5.34
5.16
30.9
30.6
1429
1425

0
65.4
30.4
35
2.61
5.52
30.6
1440

3
61.0
34.5
26.5
2.47
5.07
33.4
1404

SPECIMEN PREPARATION

Specimens tested in this study were prepared using
procedures described in the ASTM standard D5102. Initially,
the specimens were oven-dried at 60°C until the constantweight was obtained. Clay with different amounts of humic
acid was prepared by mixing relevant amounts of dry kaolinwith 0.5%, 1.5%, and 3% of humic acid by dry mass ofkaolin. Mixing of dry materials was continued until a uniform
appearance of the kaolin-humic acid mixture was obtained.
Distilled water was then added and further mixing was
performed until a homogeneous appearance of the soil paste
was achieved. This paste was then used for plasticity andcompaction tests. The specimens (76 mm in height and 38
mm in diameter) were compacted into the mould, extruded
from it and wrapped in cling film to preserve the watercontent and to keep them free from carbon dioxide (CO 2 ).
The specimens were then cured in a desiccator at 20°C andwith humidity of more than 90% for 7, 28, and 90 days,
respectively.
4.

All the specimens were cured for 7 days. It can be seen from
Fig. 1 that the undrained shear strength (Su) of the specimens
with 5% lime content increases significantly compared to thestrength of untreated clay (i.e. 0% lime). However, the shear
strength of specimens with higher than 5% lime content
reduces gradually with increasing lime content regardless ofthe humic acid content in the clay. Therefore, 5% was taken
as the optimum lime content (OLC) for each type of organic
clay. Fig. 1 also shows that the shear strength of organic clay
decreases with increasing humic acid content, which provesthat the presence of humic acid in organic clay diminishes its
shear strength.
In a further assessment of the effect of humic acid on lime
stabilisation, samples were tested at 0, 7, 28, and 90 days toascertain the duration of the stabilisation process. Since the
OLC of each specimen was identical, investigation of thedevelopment of the lime-clay reaction with time wasconducted only at 5% lime content. Fig. 2 illustrates thedevelopment of the shear strength of the lime-treatedspecimens at 0, 7, 28 and 90 days. It appears that with the
exception of the inorganic clay (0% humic acid) the
undrained strength of the lime-treated clays comprising 0.5%,
1.5% and 3.0% humic acid decreased over the 90 day curing
period. A slight loss in the strength is observed for organic
clay with 0.5% humic acid, while substantial loss in strengthis evident where the humic acid content is equal to or greaterthan 1.5%. The results demonstrate that lime stabilization oforganic clay with high humic acid content is not very efficientin the long term. Similar observations were reported by otherresearchers [8], [9], [10] and [11]. It should be noted that
despite exhibiting a loss in strength in the long term, theundrained strengths of lime-treated samples at 90 days werestill higher than those of untreated specimens (see Fig. 1 for
0% lime). However, for practical purposes, it is notreasonable to consider lime stabilisation as successful whenlong term strength is uncertain.
350

RESULTS AND DISCUSSIONS

300
250

Su (kPa)

The results of a short term strength assessment of lime-treatedclay with different humic acid contents (0%, 0.5%, 1.5%, 3%)
and lime contents (5%, 8%, 10%, 15%) are shown in Fig. 1.
300
275

150
0% humic acid
0.5% humic acid
1.5% humic acid
3% humic acid100
500% humic acid250

00.5% humic acid

225

0

10

20

30

40

50

60

70

80

90

100

Curing time (days)

200

Su (kPa)200

1.5% humic acid175

Figure 2. Effect of curing period on the undrained shear
strength of lime-treated clay.

150

3% humic acid
125
100

OL

75
50
0

5

10

15

Lime content (%)

Figure 1. Effect of lime content on the undrained shear
strength of clay with different humic acid contents after 7
days curing.

The impedance in the formation and development of
cementitious products with increasing humic acid content
over longer curing periods may be due to various reasons.
Firstly, the reduction in pH value at longer curing periods
may be one or the reason that decreased the shear strength
value. Fig. 3 shows the pH value of various humic acidcontents treated at OLC = 5% for different curing periods. It

254

12.9

humic acid content without the addition of salt. The changes
in soil structure due to lime stabilisation were hardly visible.
The microstructure of the specimen was very similar to that ofa specimen without lime [21].
600500

Su (kPa)

can be seen that, except for inorganic clay, the pH values oforganic clays decreased with increasing curing period. After
28 days, specimens which consist of 1.5% and 3.0% humic
acid contents had a pH lower than 12.40, which is theminimum value required for the successful production of
cementing materials [14].
Secondly, bearing in mind that the dissolution of clay
minerals is highly dependent on pH value, it is thought thatthe presence of a certain amount of humic acid made itdifficult for the solution to recover from its acidified
condition. The humic acid may have coated the clay particles,
thereby preventing the lime from coming into contact withclay minerals during the pozzolanic reaction. Consequently,
the dissolution of clay minerals becomes insufficient, thus
limiting the production of cementitious materials.

400
0.5% CaCl2
0.5% NaCl
0% salt

300
200
100
0

0

10

20

30 40 50 60 70
Curing time (days)

80

90

100

Figure 4. Effect of the chloride salts on the shear strength oflime-treated clay at different curing periods.

12.8
12.7

pH

12.6
12.5
12.4
12.3

0% humic acid
0.5% humic acid12.2

1.5% humic acid12.1

3% humic acid

12.0

0

10

20

30

40

50

60

70

80

90 100Curing time (days)

Figure 3. Effect of curing period on pH value of lime-treatedorganic clay with different humic acid contents.
The findings obtained from the experiments on lime-treatedorganic clay revealed that lime may not be suitable forstabilising organic clay with more than 1.5% humic acidcontent [15]. Therefore, two chloride salts (calcium chloride(CaCl2) and sodium chloride (NaCl)) were considered as
potential additives that could improve the effectiveness oflime stabilisation in organic clay. The analysis was carriedout for the clay with 1.5% humic acid content stabilised with
5% of lime.
Fig. 4 shows development of the undrained strength withincreasing salt (CaCl2 and NaCl) content for different curing
periods. The salts-treated specimens are also compared tothose treated by lime alone (i.e. 0% salt content) atcorresponding curing periods. A clear increase in the strength
can be noted for both types of salt as the curing period
increases from 7 to 90 days. Overall, the strengths recorded
for the lime-CaCl 2 mixtures were higher than those of thelime-NaCl mixtures [13]. A similar finding was reported by[12]. It is may be because of the dual function of Ca2+ ionspromoting better ion exchange and contributing towards theformation of cementing products (i.e. calcium silicate
hydrate, CSH and calcium aluminate hydrate, CAH). The Na+
ions, on the other hand are only believed to accelerate silica
dissolution [9], [12] and [13].
Changes in soil structure and chemical compositions wereanalysed using a Scanning Electron Microscope (SEM). Fig.
5 shows the image of a lime-treated specimen with 1.5%

Figure 5. SEM image of lime-treated clay with 1.5% humic
acid and without the addition of salt.

Figure 6. SEM image of lime-treated clay with the addition of
0.5% CaCl2 at 28 days of curing.

255

4.

The SEM images of lime-treated organic clay show theflocculated and aggregated structure, without appearanceof cementation. In contrast, a clear evidence of cemented
structure is observed when lime-treated organic samples
are stabilised by addition of 0.5% CaCl 2 and 0.5% NaCl.

6. REFERENCES

Figure 7. SEM image of lime-treated clay with the addition of
0.5% NaCl at 28 days of curing.
Fig. 6 and Fig. 7 show the SEM images of lime-treatedorganic clay with the addition of 0.5% chloride salts. The
micrographs were taken from UCS test samples after 28
curing days. The flaky and plate-like structure present in thelime-treated organic specimen (Fig. 5) has almost completely
disappeared. A clear cemented structure, mainly calcium
silicate hydrate (CSH) was identified by a fine tubular shape
and well knit structure (needle-like form) that bridge theaggregates [13], [15]. The presence of cementing structures in
specimens with the addition of salts explains the
improvement in the strength obtained from UCS tests.
5. CONCLUSIONSIn this paper, the effectiveness of lime stabilisation of organic
clay was investigated. In addition, a preliminary investigationon the effect of salts on the strength of lime-treated organic
clay was analysed. Based on the experimental results
obtained in the study, the following conclusions can be made.
1. The shear strength of lime-treated organic clay reduces
when the lime content exceeds 5%. Thus, 5% of lime isidentified as the optimum lime content (OLC) for theorganic clay tested in this study.
2. The shear strength of the lime-treated organic clay
reduces at longer curing periods. A slight loss in strengthis observed for organic clay with 0.5% humic acid, while
a substantial loss in strength is evident where the humic
acid content is equal to or greater than 1.5%. These
results show that the presence of more than 1.5% humic
acid in the organic clay tested in this study reduces
significantly the efficiency of the lime stabilisation
process.
3. The results obtained from the UCS tests carried out on
specimens with 1.5% of humic acid treated with 5% lime
show that the strength of lime-treated organic clay can be
successfully improved with addition of 0.5% chloride
salts. Therefore, there is a potential for the use of chloride
salts in lime stabilisation of organic clay in geotechnicalpractice.

[1] Bell FG, “Lime stabilization of clay minerals and soils,”
Engineering Geology, 1996, vol. 42, no. 4, pp. 223-237.
[2] Rajasekaran G and Rao N, “The microstructure oflime-stabilized marine clay,” Ocean Engineering, 1997,
vol. 24, no. 9, pp. 867-875.
[3] James R, Kamruzzaman AHM, Haque A, Wilkinson A,“Behaviour of lime-slag-treated clay,” Proc. the ICE,
Ground Improvement, 2008, vol. 161, no. 4, pp. 207-216.
[4] Consoli NC, da Silva Lopes Jr L, and Heineck KS, “Keyparameters for the strength control of lime stabilizedsoils,” J. of Materials in Civil Engineering, 2009, vol. 21,
no. 5, pp. 210-216.
[5] Kuno G, Kutara K and Miki H, “Chemical stabilization
on soft soils containing humic acid,” Proc. 12th
International Conference Soil Mechanics Foundation
Engineering, 1989, Rio de Janeiro, vol. 2, pp. 1381-1384.
[6] Hebib S. and Farrell ER, “Some experiences on thestabilization of Irish peats,” Canadian Geotechnical J.,
2003, vol. 40, pp. 107-120.
[7] Chen D, Si Z and Saribudak M, “Roadway heaving
caused by high organic matter,” J. of Performance ofConstructed Facilities, 2009, vol. 23, no. 2, pp. 100-108[8] Huat BB, Maail KS and Ahmed Mohamed T, “Effect ofchemical admixtures on the engineering properties oftropical peat soils,” American J. of Applied Sciences,
2005, vol. 2, no. 7, pp. 1113-1120.
[9] Koslanant S, Onitsuka K and Negami T, "Influence ofsalt additive in lime stabilization on organic clay,”
Geotechnical Engineering, 2006, vol. 39, pp.95-101.
[10] Harris P, Harvey O, Sebesta S, Chikyala SR, Puppala A,
and Saride S, “Mitigating the effects of organics instabilized soil,” Technical Report No. 0-5540-1, 2009,
Texas Transportation Institute, USA.
[11] Zhu W, Chiu CF, Zhang CL and Zeng K, “Effect ofhumic acid on the behaviour of solidified dredged
material,” Canadian Geotechnical J., 2009, vol. 46, pp.
1093-1099.
[12] Modmoltin C and Voottipruex P, “Influence of salts onstrength of cement-treated clays,” Proc. the ICE, Ground
Improvement, 2009, vol. 162, no. 1, pp. 15-26.
[13] Danmarks B and Allan Haliburton T, “Effects of sodium
chloride and sodium chloride-lime admixtures oncohesive Oklahama soils,” Highway Research Record,
Washington, 1999, vol. 315, pp.102-111.
[14] Mohd Yunus NZ, Wanatowski D, and Stace LR, “Effectof humic acid on physical and engineering properties oflime-treated organic clay,” Proc. World Academy of
Science Engineering and Technology, 2011, Issue 59,
Part XV, 1820-1825.
[15] Mohd Yunus NZ, Behaviour of lime-stabilised organic
clay. PhD Thesis, 2011, University of Nottingham, UK.

256

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia,Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Sodium Tripolyphosphate For Stabilizing Of Contaminated Soil
Amin Falamaki1,HosseinTavallali2 and ShahrbanooRezanejadFarahmand 3
123
Payame Noor University, PO BOX 19395-3697 Tehran, Iran
P

P

P

P

P

P

P

ABSTRACTThe objective of this study is to stabilize the polluted soil using sodium tripolyphosphate (STPP). In this study a seriesof leachate column tests were conducted to find out the effect of STPP on Pb polluted soil. In addition, to understand theconcentration of STPP on the efficiency of the method, different samples with different polluted soils were tested. STPP
with 0.1, 0.2 and 0.5 mg/kg dry soil were added to the polluted soil and the samples were kept for 1 month. The results
show that STPP may stabilize Pb in the soil. Increasing the concentration of STPP, decreases the concentration of
metals in the effluent, means more stabilized metals in the soils. The results also show that 0.2 mg/kg drysoil of STPP isenough to stabilize the metals from the first stages of the tests.
Keywords: Lead Stabilization, Phosphate amendment,Contaminated soil.

1. INTRODUCTION
Today in developing countries, unfortunately, due to lack ofadvanced technologies, disposal of contaminated soils withheavy metals and radioactive in the environment, causes
spread of contaminants in groundwater. In addition, methods
for decontamination of the soil are very expensive and require
advanced technology. Therefore, disposal of mining waste
and industrial soil waste using an inexpensive method thatprevents from spreading pollution of soil is very important.
Solidification/stabilization technique is relatively a simple
and inexpensive method for removing contaminants from thesoil. In this method one stabilizer is used to immobilize
pollutants in soil and prevent them from spreading over. This
technique is especially useful for urgent disposal ofcontaminated soils in the environment.
Lee et al., (2009) stated different amendments that used insitu stabilization of heavy metals in soils, including
agricultural products, such as lime, phosphate, and organic
matter, as well as various industrial products, such as zeolites
and red-mud. [1]Falamaki et al., (2009) declared the phosphate based
binders have many applications in geotechnical andgeoenvironmental engineering such as: improvement andstabilization of tailing dams, rehabilitation of old landfills,
construction of new clay liners, containment and permeablebarriers, etc. [2]Mignardi et al.,(2011) showed, The reaction mechanisms
for metal immobilization by phosphate minerals which
includes: (a) ion exchange process, (b) surface complexation,(c) dissolution of the original phosphate minerals andprecipitation of new metal phosphates, and (d) substitution of
Ca in phosphate by other metals during recrystallization(coprecipitation). [3]Orthophosphate (PO 4 3-) is one of metals stabilization
R

RP

P

agents that have many applications to stabilize a variety ofhazardous and industrial wastes. This agent combines withmetals to form phosphate minerals, and reducing solubility
metals through absorption or precipitation and immobilize
them. [4]Laperche et al.,(1996) declared The use of phosphate toimmobilize metals has been advocated for lead-contaminatedsoils.[5]The purpose of this research, is stabilization of lead (Pb) incontaminated soil , using STPP in three stabilizer / soil ratio(0.5/100, 0.2/100, 0.1/100 w/w) as a chemical immobilization
agent, using leachate column test described in ASTM D
4874- 95. The optimum amount of STPP is also investigatedin this paper.
2. MATERIALS AND METHODS
2.1 Soil description
A fine silty sand from Shiraz sand mines (Fars Province, Iran)was used in this study. The amount of soil moisture content isequal to 0/94%. Soil pH is 7. The aggregation analysis using
sieve showed that 57.4% was sand.
2.2 STPP characterization
One of the phosphate compounds that used in this study is
sodium tripolyphosphate, STPP, A white powder, which is
inexpensive. This detergent mater forms complexes withwater ions, such as calcium and magnesium. The other uses of
STPP are emulsifier of oil and grease, peptizing agent,deflocculating agent in oil well and sequester in cotton
boiling (Tavallali et al., 2010)
2.3 Leachate column test
Apparatus that used in this study were Sens AA flame atomic
absorption spectrometer. Standard solution of (lead (II)nitrate) and stabilizer (sodium tripolyphosphate) obtained
from Merck as analytical materials and were used without
further purification. All dilute solutions were prepared from257

double distilled water.
Experimental program is based on soil column experiment.
Soil samples are prepared according to leachate column test.
This is a standard method described in ASTM D 4874-95 isfully. In order to achieve the research objectives, soil samples
were dried by oven at temperature= 150° C for 24 hours. The
spiked dry soil sample was prepared by mixing the soil with
Pb(NO 3 ) 2 solution at a concentration of 500 mg Pb/kg dry
soil, and were cured for one month to prevent loss of moisture.
More investigations should be accomplished to determine theoptimum and/or most economical content for the different
purposes such as contaminant stabilization. The ratios of the
STPP/dry soil are chosen as 0.1, 0.2, and 0.5 % in this study.
STPP -dry soil mixtures were kept for 2 week. After 2 weeks,then the leachate column tests were conducted. Also
measurements were done at pH≈5, (using HCl 0.1M), androom temperature

Influent
Tank 1
Valve 1
Tank 2h=1m

Valve 2
Teflon filter and paper filter
Soil column
Valve 3Effluent
Fig. 1. Schematic diagram of the apparatus
The leachate column test is a standard method described in Figure 1 illustrates a schematic diagram of the apparatus
ASTM D 4874-95 is fully. The Leachate column test was which is designed for this study. Comparison of data obtained
carried out by passing the leachate through a bed of dry soil or from the latex output using atomic absorption spectroscopy
soil-contaminant (Pb) mixtures or soil- contaminant apparatus was used to evaluate the efficiency of the method.
(Pb)-stabilizer mixtures contained in a column. Passing the
2.4 EXPERIMENTAL PROGRAM
leachate is conducted in an up-flow mode. Column (56.4mm
in diameter and 74.6 mm in height), was uniformly packed To achieve the objectives of this research an experimental
with soil samples. Maximum amount of soil poured and program is scheduled as given in Table 2.
compacted in each cell was 280 gr.
Table 2.EXPERIMENTAL PROGRAM
Row

Experiment

STPP/soilratio

1

T0 - No additive- No pollute-

pollution and stabilizer
R-

Storage time forpolluted sample

curing time for polluted
sample and stabilizer

-

-

R

2

T1- No additive- Add pollute

0/100

Only pollution

1 month-3

T2-Additive-contaminated soil

0.1/100

pollution then stabilizer

1 month

2 week

4

T3-Additive-contaminated soil

0.2/100

pollution then stabilizer

1 month

2 week

5

T4-Additive-contaminated soil

0.5/100

pollution then stabilizer

1 month

2 weekVoid ratio (e) for columns with varied stabilizer / soil ratios

calculated and results are shown in Table. 3. In this method258

for evolution of efficiency of columns with varied stabilizer /soil,all columns are similar in void ratio to provide

SamplesT0 until T7

comparability of the samples.

Table 3. Properties of samples
STPP/soil
Dry soil
γderatio
weight(g)

0/100,0.1/100,0/2/100and 0/5/100

280

3. Results and Discussion
To interpret the leachate column test results break throughcurves for each test in Table. 2 were plotted. As depicted inTable. 2 a control test (T0) was conducted to show that the

1.50

0.76

n

pore
volume

0.431

80.25

amount of the Pb is negligible in the soil. Results obtained
from T0 in Fig. 2 shows that soil sample is clean and without
any contaminant.

Fig.2. Control test (T0), soil sample with no pollutant and additiveFig.3..Pb contaminated soil sample (T1) with no additive259

After polluting the soil with Pb, the concentration ofcontaminant in the effluent vs pore volume wasplotted in Fig. 3 and it is seen that as the clean water
passes through the soil, Pb leachates continually bywater. Fig. 4 shows the effect of STPP stabilizer on
the polluted soil. According to this figure, adding

STPP to the polluted soils decreases the extracted
Pb in the effluent water. Increasing the amount of
STPP, decreases the extracted Pb, this means more
immobilization of Pb. However, the optimum
amount of STPP seems to be 0.2% by weight for thissoil condition.

Fig.4.effect of STPP stabilizer on the polluted soil

[4] Tavallali H, Falamaki A, Salehi H, 2010.
“Stabilization of Metal contaminants in
Municipal Waste Leachate using Metal
precipitant additives.,”J. ofChemtech Aug.
2010,pp. 1-2[5] Laperche, V., Traina, S.J., Gaddam, P., Logan,
T.J. “Chemical mineralogical characterizationof Pb in a contaminated soil: reactions withsynthetic apatite.” Environ. Sci. Technol. 1996,
30, 3321–3326.

4. Conclusion
This study investigated the effectiveness of STPP in
immobilization Pb in fine silty sand.
All results showed that complexation reaction andstabilization process occur in any ratio of STPP.
Increasing the concentration of STPP, decreases theconcentration of metals in the effluent, means morestabilized metals in the soils. The results also showthat 0.2 mg/kg dry soil of STPP is enough tostabilize the metals from the first stages of the tests.
5. REFERENCES[1] Lee S.H, Lee J.S, Choi Y.J, Kim J.G, “In situ
stabilization of cadmium,lead, andzinc-contaminatedsoilusing
variousamendments,”J. Chemosphere 77Aug. 2009,
pp. 1.
[2] Falamaki a, Shariatmadari N, Noorzad
A,“Strength Properties of Hexametaphosphate
Treated Soils,” J. of GEOTECHNICAL AND
GEOENVIRONMENTAL ENGINEERING
Aug. 2008,pp. 1. ,ASCE. 1090-0241.
[3] Mignardi S, Corami A, Ferrini V, “Evaluationof the effectiveness of phosphate treatment fortheremediationof
mine
waste
soils
contaminated with Cd, Cu, Pb, and Zn,”Aug.
2011, pp. 1.
260

261

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Evaluation of Soil Water Characteristic Curves of Sand-Clay Mixtures
T. Y. Elkady1, M. A. Dafalla1, A. M. Al-Mahbashi1 and Mosleh Al Shamrani1
1King Saud University, Saudi Arabia
ABSTRACTThis paper presents an experimental investigation performed to evaluate the soil water characteristic curves ofsand-Al-Qatif clay mixtures. Al-Qatif clay is natural expansive clay widely spread in the eastern province of Saudi
Arabia. Detailed mineralogical and physical tests were performed on Al-Qatif clay for the characterization. Mixturesof sand and Al-Qatif clay were prepared with sand being the base material and with different clay contents (0%, 5%,
10% and 15%). Test results indicate that the shape of SWCC curves changes from the unimodel to the bimodal form ofsoil water characteristic curve with the increase in clay content indicative of changes in sand/clay mixture micro-andmacro-pores.
Keywords: Soil Water Characteristic Curves, Sand- clay mixture, soil suction

1. INTRODUCTION
Hydraulic barriers are integral components of various
engineering systems such as waste containment facilities,
earth dam, and water conservation practices. The purpose of
these barriers is to preclude the migration of water orleachate. Compacted sand-bentonite barriers have proven to
be a suitable material for hydraulic barriers. Several
researchers [1]-[6] reported that the amendment of sand withbentonite (ranging between 4 and 20 percent) yielded
hydraulic conductivity values acceptable to engineering
standards. In addition, sand mixed with bentonite yielded acomposite material with satisfactory characteristics in regards
to shear strength, susceptibility to desiccation cracking, andease of construction.
The Kingdom of Saudi Arabia (KSA) is one of thecountries which have considerable areas of expansive soil
formations. These formations cover vast area of KSA
estimated to be about 800,000 km2 [7] (Fig. 1). This study
evaluates the hydraulic characteristics of compacted
sand/Al-Qatif expansive clay hydraulic barriers. Al-Qatif
clay represents a natural expansive clay that is abundant in the
Arabian Gulf coastal region (Fig.1). This over-consolidated
clay is extremely fissured in the natural unsaturated state andexhibits significant volume changes when water content isaltered. The volume change potential is due to the highamount of expansive clay minerals such as smectite and illite[8].
The hydraulic characteristics and proper simulation ofwater flow through compacted sand-clay barriers requireconsideration of unsaturated liquid flow principles. One of
the main unsaturated soil parameters required for modeling is
the soil water characteristic curve (SWCC). SWCC is definedas the relationship between soil suction and water content(whether gravimetric, volumetric).

Fig. 1 Distribution of expansive soil formations in Saudi
Arabia [7].
Previous studies [9]-[11] focused on the evaluation of
SWCC of compacted sand-bentonite mixtures. Results ofthese studies revealed that the retention capacity increased
with increase in bentonite content. Reference [9] indicatedthat the SWCC possessed a hysteresis effect depending on thewetting and drying paths. In addition, the initial density had amarkable effect on the air entry value of SWCC.
Review of technical literature showed a possible trend
regarding the shape of SWCCs of sand-bentonite mixtures.
SWCCs with unimodel shape were reported forsand-bentonite mixtures with high bentonite content (greaterthan 30%) [10]-[12]. However, SWCC with bimodal form
were observed in case of low bentonite content-sandmixtures. References [13], [14] reported that the SWCC ofsand-bentonite mixture with 8% bentonite content revealed abimodal shape (i.e., curve with four bending curves). This261

was attributed to pore water trapped between both the larger,
interaggregate pores between the sand particles and bentonite
platelets and the smaller, intra-aggregate pores.
This paper presents aims at performed evaluating theeffect of Al-Qatif clay percentage on the SWCCs ofsand/Al-Qatif clay mixtures. Clay content considered were0%, 5%, 10%, and 15% per dry mass mix. This was achievedby measuring the soil water characteristic curves over a rangethat captures the main features of the curves.

depicts that Al-Qatif clay consist of montmorillonite andpalygorsite which are typical swelling minerals.
Uniform sand was used in this study with grain size
ranging from 0.6 to 0.1 mm. The grain size distribution ofsand used is shown in Figure 2. According to unified soil
classification system, the sand was classified as poorly gradedsand (SP).

2. MATERIAL AND METHODSThis section describes the material used for the compacted
sand-Al-Qatif expansive clay mixtures.
Initially, thegeotechnical properties of each of the sand and clay material
are presented. Finally, the effect of clay content on thegeotechnical characteristics of sand clay mixture ishighlighted.

Table 1. Soil Characterization Data for Al-Qatif Soil________________________________________________
Test
Value
________________________________________________
2.70
Specific Gravity, G s
Liquid Limit, w L (%)
137 %Plastic Limit, w P (%)
60 %Shrinkage Limit, w sh (%)
12 %% passing Sieve No. 200
99 %Unified soil classification
CH
Standard Proctor Compaction Test
Maximum dry unit weight (kN/m3)
12 kN/m3
Optimum water content (%)
40 %Fig. 1. X-Ray diffraction analysis.

100

Red Sand Analysis

90

Percent finer (%)

2.1 Soil Used
The expansive clay used in this study was obtained from thecity of Al-Qatif located on the Arabian Gulf shoreline at a
distance of 400 km from Riyadh the capital of Saudi Arabia.
Several researchers investigated the swelling characteristics
of Al-Qatif expansive clay [8], [15], [16]. Based on theseinvestigations, Al-Qatif clay was characterized as highly
expansive soil due to the presence of high montomorillonite
mineral content.
Soil samples were obtained from open pits excavated to adepth of 1.5 - 3.0 m below ground surface. Samples weretransferred to laboratory and complete mineralogical andgeotechnical characterization was performed. A summary ofgeotechnical characterization results is presented in Table 1.

80
70
60
50
40
30
20
10
0

100

10
1
0.1
Grain size Diameter (mm)0.01

Fig. 2 Grain size distribution of sand.

Swelling potential (ASTM D4546)
16-18%
Swelling pressure (ASTM D4546)
550-600 kN/m2________________________________________________
Mineralogical characterization of Al-Qatif soil was
performed using Bruker D8 Advance system. Samples were
scanned from 2° to 60° (2θ) using 2.2kW Cu anode long fine
focus ceramic X-ray tube at a scanning rate of 1 degree perminute. The X-Ray diffraction analysis shown in Figure. 1

2.2 Compaction Curves of Sand-Clay Mixtures:Standard compaction tests were performed in accordancewith ASTM D 698 [17] to evaluate the compaction
characteristics of sand-clay mixtures. The results of the tests
are presented in Fig. 3 where it is observed that maximum
unit weight increased with increase of clay content, while theoptimum water content increased with increase in clay
content.

262

18.5018.00
Dry unit weight for Soil (kN/m3)

using a balance with a 0.0001 resolution. At the end of thetest, the final water content of the sample was calculatedusing gravimetric methods. The water content at each suctionlevel was back-calculated based on the final water content
and the amount of water loss recorded at each suction level.
The hanging column apparatus comprise of a glass Buchner
funnel with a porous plate and two graduated burette system
connected to the outflow end of the funnel as shown in Fig. 4.
The purpose of these burettes is to apply suction and measurewater outflow during the test. The "suction application"
burette contains a valve to seal the top of burette from theatmosphere. Suction was applied by raising the Buchner
funnel such that base of the sample is above the water level inthe suction application burette when the valve is closed.
Suction is calculated as the difference between water level
multiplied by unit weight of water. The "water outflow"
burette was used to monitor and record the amount of water
expelled or entered during the test. Equilibrium was attained
when no change in water level was observed. Water content
at each suction level was calculated based on final watercontent of sample and amount of water expelled at eachsuction level.

Clay Percent
0%
5%
10%
15%

17.50
17.00
16.50
16.00
15.50
15.00

0.0

5.0

10.0
15.0
Water Content %

20.0

25.0

Fig. 3 Compaction curves for sand-clay mixtures withdifferent clay contents
2.3 Sample Preparation
Samples of Al-Qatif expansive clay obtained from the field
were air dried, pulverized and sieved using sieve No. 40. The
sand and clay content were mixed thoroughly and then
optimum water content corresponding to each mix was added
and stored in plastic bags for 24 hours to mellow. Compacted
samples of sand- AlQatif clay mixtures were prepared for thedetermination of the soil water characteristic curves. Samples
were statically compacted in a stainless steel ring tomaximum unit weight corresponding to that presented in Fig.
3. All samples were 50 mm in diameter and 20 mm thick.
3 EXPERIMENTAL PROGRAM
2.2 Soil water characteristic Curves
The drying curve of the SWCCs of sand- AlQatif clay
mixtures were evaluated using pressure plate test and hanging
column technique. Pressure plate extraction apparatus isbased on the principle of axis translation technique [18] andwas used to apply suction ranging from 5 kPa to 1500 kPa.
The hanging column technique was used to evaluate the
SWCC for the suction levels less than 5 kPa. All tests were
performed according to ASTM D6836 [19].
For the pressure plate extractor test, samples were placedin confined compartment with a ceramic disc base of air entry
value of 5 or 15 bars. Prior to the test, the ceramic disk was
saturated by soaking it in water of 24 hours. Suction on
samples was applied in increments and water exiting from thecompartment below the ceramic disc was monitored.
Equilibrium under each suction level was assumed to beattained when no water egress was observed. Afterequilibrium under each suction level, the amount of water lostfrom tests samples was determined by weighing the samples

Fig. 4 Hanging column apparatus
4 RESULTS AND DISCUSSIONSThe SWCCs for sand-expansive clay mixtures under different
clay contents are depicted in Fig. 5. In Fig. 5, a gradualtransition from a unimodel SWCC to a bimodal SWCC wasobserved as clay content increases. The unimodel SWCC is263

characterized by having two bends in the curve defining the
air entry value and residual water content (Fig. 5). The air
entry value is defined as the matric suction above which aircommence to enter the soil pores. The residual water content
is defined as the water content beyond which no significant
decrease in water content occurs. The bimodal SWCC ischaracterized by having four distinct bendings: two air entry
values and two residual water contents. For clay content equal
to or less than 5 percent, the SWCC show a unimodal form of
SWCC. With the increase of clay content greater than 5%,
the SWCC show a bimodal form.
It is further observed from Fig. 5 that the residual watercontent increases with the increase of the clay content. In
addition, the air entry value increases with the increase in clay
content. These observations are attributed to the presence ofsmaller pore size developed as a result of clay particles fillingthe voids between sand particles.
Bimodal SWCC are generally observed for gap-grade grain
size distribution soils [20] as well as soils that includes twolevels of pore sizes defined as macropores and micropores[21]. Therefore, it can be inferred that the increase of clay
content resulted in the formation of mesopores/micropores
within the compacted sand-clay mixtures. The portion of thesoil water characteristic curves that represent macropore size
ranges from matric suction 0.1 to 100 kPa. Whereas, theportion of the SWCC representing mesopores/micropore
sizes lies between matric suction 200 and 1500 kPa.

Gravimetric water content (%)

25.0

Clay Content
0%
5%
10%
15%

20.0
15.0
10.0
5.0
0.0
0.01

0.1

1

10

100

Matric Suction (kPa)

1000

10000

Fig. 5 Soil water characteristic curves of sand Clay Mixtures
5 SUMMARY AND CONCLUSIONSIn this study, the effect of clay content on the soil water
characteristic curves of compacted sand-AlQatif clay
mixtures were evaluated. Clay contents considered include
0%, 5%, 10%, 15%. Results from this study indicated that, as
the clay content increased, the shape of the SWCC ofsand-clay mixtures transforms from a unimodal form to abimodal form.
Furthermore, the residual water content was observed toincrease with increase in clay content signifying increase inwater retention capacity increases with increase in clay
content.
The bimodal form of the SWCC indicates the presence of
two levels of pore sizes; namely macropores andmesopores/micropores. For 10% and 15% clay content, themacropores are considered the dominant pore size covering abroad range of the SWCC from 0.1 kPa to 100 kPa. Therefore,
it is inferred that the SWCC of compacted sand-clay mixturesare strongly related to the texture and pore size distribution of
the sand-clay mixture which in turn, has a significant impact
on its hydraulic characteristics.
6 ACKNOWLEDGMENT
This paper is a part of a research project supported through
NPST program by King Saud University, Project No. ENV
1183. The authors thank the staff of Bugshan Research Chair
in Expansive Soils and especially Eng. Abdulla Shaker for his
works in constructing the soil compaction curves for sandclay mixtures.
7 REFERENCES[1] Cho. WJ, Lee, JO, and Kang, CH,. "Hydraulicconductivity of compacted soil-bentonite mixture for aliner material in landfill facilities," Korea Society ofEnvironmental Engineers, Environ.Eng. Res. Vol. 7. No.
3, 2002, pp. 121-127.
[2] Komine, H, “Simplified evaluation on hydraulicconductivities of sand–bentonite mixture backfill,”
Applied Clay Science , vol. 26, 2004, pp. 13– 19[3] Chalermyanont, T. and Arrykul, S, "Compacted
sand-bentonite mixtures for hydraulic containment
liners,” J. Sci. Technololgy,Vol. 27,No.2, Mar - Apr
2005.
[4] Ameta, NK and Abhay, S W, "Effect of bentonite on
permeability of dune sand," EJGE, Vol 13, (2008).
[5] Akgün, H, “Geotechnical characterization andperformance assessment of bentonite/sand mixtures forunderground waste repository sealing,” Applied Clay
Science, Vol. 49, 2010, pp. 394–399.
[6] Gueddouda,. MK, Lamara, MNA, and Taibi, S,"Hydraulic conductivity and shear strength of compacted
dune sand - bentonite mixtures,” ICCBT, 2008.
[7] Ruwaih, IA, “Experiences with Expansive Soils in Saudi
Arabia,” in Proc. 6th Int. Conf. on Expansive Soils, New
Delhi, India, 1987, pp.317-322.
[8] Azam, S, "Influence of mineralogy on swelling andconsolidation of soils in eastern Saudi Arabia," Canadian
Geotechnical Journal, Vol.40, No. 5, 2003, pp. 964–975.
[9] Montanez, JEC. "Suction and volume changes ofcompacted sand-bentonite mixtures," Ph.D.
Dissertation, Imperial College of Science, 2002
[10] Agus,SS and Schanz,T, "wetting, drying, suction
characteristic curves of bentonite sand," Proceedings ofthe Fourth International Conference on Unsaturated
Soils, April 2-6, 2006, Carefree, AZ, pp. 1405-1414.

264

[11] Pei-yong, L and Qing, Y, "Test study on soil-water
characteristic curve of bentonite-sand mixtures," EJGE,
Vol. 14, 2009, pp. 1-8[12] Sun, D et al, "Hydro-mechanical behaviors of highly
compacted sand-bentonite mixture," Journal of RockMechanics and Geotechnical Engineering, Vol. 2, No. 1,
2010, pp. 79–85.
[13] Stoicescu, J T, " Soil water characteristics and pore size
distribution of a sand-bentonite mixture," In Proc. 49th
Canadian Geotechnical Conference, St. John's
Newfoundland, September 23-25, Vol 2, 1996, pp.
721-728.
[14] Stoicescu, JT et al., "Soil water characteristics ofsand-bentonite mixtures used for liners and covers,"
Second International Conference of Unsaturated Soils,
Beijing, China, August 27-30, Vol 1, 1998, pp. 143-148.
[15] Abduljuawad, SN, Hameed, RA, Al-Sulaimani, GJ,
Basunbal, IA, Safar, MM, “Expansive soils in Eastern
Province of Saudi Arabia,” Proceedings of 7thInternational Conference on Expansive Soils, Dallas,
TX, vol. 1, 1992, pp. 426-431.
[16] Azam, S, Abduljauwad, SN, Al-Shayea, NA,
Al-Amoudi, OSB, “Expansive characteristics ofgypsiferous/anhydritic formations,” Engineering
Geology, Vol. 51, 1998, pp. 89-107.
[17] ASTM D698 - 02 Standard Test Methods Standard Test
Methods for Laboratory Compaction Characteristics ofSoil Using Standard Effort (12 400 ft lbf/ft3 (600 kNm/m3) , Vol. 15.09, Annual book of ASTM Standards,
West Conshohocken, PA.
[18] Hilf, JW, "An investigation of pore-water pressure incompacted cohesive soils," Ph.D. Thesis, Tech. MemoNo. 654, U.S. Dept. of interior Bureau of Reclamation,
Design and Construction Div., Denver, Colorado, 1956
[19] ASTM D6836 - 02 Standard Test Methods for
Determination of the Soil Water Characteristic Curve for
Desorption Using a Hanging Column, Pressure
Extractor, Chilled Mirror Hygrometer, and/orCentrifuge, Vol. 15.09, Annual book of ASTM
Standards, West Conshohocken, PA.
[20] Durner, W, "Hydraulic conductivity estimation for soils
with heterogeneous Pore Structure," Water Resources
Research, Vol. 30, No. 2, 1994, pp. 211-223.
[21] Burger, CA and Shackelford, CD, "Soil-Water
Characteristic Curves And Dual Porosity ofSand–Diatomaceous Earth Mixtures. Journal Of
Geotechnical And Geoenvironmental Engineering, Vol.
127 No. 9, September 2001, pp. 790-800265Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Coastal soils improvement
Mohammad reza Atrchian 1, Morteza Ala2 , Amin Naeimabadi 3
Islamic Azad University Of Zanjan, Iran
P

P

P

P

P

ABSTRACTwith an ever-increasing population growth, the development of industries and the deficit problem of land forestablishing the required constructions have been caused thatpeopleuse the lands with lower engineering

qualification . But for establishing a suitable construction the qualification of soil should be improved by a technique
respect to engineering parameters. In this condition it has a suitable operation in effect of involving forces. Soil
improvement techniques can be classified in different points of view. These methods are divided in to three generalgroups: Compaction, Solidification and Stabilization. With regard to this point that coastal soils have the capability ofliquefaction and they are clayey soils, in this article, first of all we will inspect these soils and the problems that they arecaused and then their characteristics improvement techniques based on geotechnical methods such as placing trenches
for limited land, sand wills for liquefiable, saturated and clayey soil ;stone columns for clayey soil that have bearing
capacity problem and dual purpose micro pile and compaction will be surveyed. Outdoor investigations and theirfindings show that the soils with SPT lower than 12 in each depth are problematic surely and should improved.
Keywords: soils improvement, liquefaction, stone column, micropiles, compaction

consideration of pre-acting construction but also of reword

1. INTRODUCTION
Soil structure (construction) can be considered as an ideal andsuitable place by engineers to have 4 features. 1. Having
enough shear strength and bearing capacity of the soil. 2.
having low degree of immediate settlement and consolidation
caused by the load. 3. Acceptable changes of volume
expansion of the soil (for instance, swelling caused byunloading or humidity rise in clay soil) or volume contractionof the soil (caused by decreased humidity) so that theconstruction effacing won’t be conflict. 4. Not having any
serious problem in construction place.
Houseman (1990) has basically classified soil improvement
goals as the following.
Soil bearing capacity increase, soil settlement decrease, void
decrease and soil density increase, regularizing of the featuresof changeable materials. Improvement requires not only

and the meantime construction. There are some major
methods like compression and micropile, but according to themodern technology, new ways of biotechnology has beenfounded that can prepare the life condition for undergroundanimate to fill up the holes of the sand that prevents waterpressure while earthquake and finally there won’t be soilliquefaction. Or for the soil dams the hales can be filled bybacteria injection and rising them instead of providing therequired amount of water at the centre.
2-Based on boring log there are two kinds of problematicsoil
One kind is soft clay soils and the other kind is susceptibleliquefaction soils. In deeper areas there are smaller sands thatare susceptible for liquefaction, but coast line sands are more
average & bigger. The more stickiness the more resistance ofthe sand against liquefaction.

266

2-1. Liquefaction is a phenomenon that is resulted of a2-2. According to the experiments on the coastal soil

decrease in hardness & resistance of the soil caused byconsequences (13, 14 and 15 SPT ) there is a doubt of theearthquake and quick loading. This phenomenon and its

being liquefiable soil, but SPT lower than 12 in any depth willconsequences mode huge damages in recent years around thecertainly be problematic. Capable liquefaction soil can beworld. Which usually happens in saturated soils. Existingbuilt by silt rather than fine grained sand but remember thatwater in soil particles gets soil particles under pressure andthe mixture of silt an fine grained sands is really qualified forthis event causes strong compression of the soil. Before theliquefaction that the cleaner and through water path.

earthquake water pressure in not considerable, but after
starting it water pressure increases that the result is soil

3-geotechnical improvement method based on the coast

particle movement. However the mentioned factor can beline areas:reason for the liquefaction phenomenon. There are also ether

3-1: trenches: in small land where we have just 2 or 3 meterreasons like constructing activating such as explosion andsoil for improving, if we want to protect our own building,basically any changes in the ground by loading andwell do trenches around the land. For a 30 cm thick sealing in

unloading.

the case of using bentonite with 4% cement, there will be atime of 3years for water to pass just 1cm. so for passing about
30 cm around 90 years will be needed.
3-2. Compaction methods:
In rural areas and around the built building we can’t roller on
the other hand the action of compaction can’t be used incoastal area sands more than 30 - 40 cm. dynamic compaction
method will be suitable for liquefiable soil and loose grainthat have the problem of bearing capacity and fine grain soil.
Disadvantages of this method are noise pollution not being

Fig 1 : water ,saturated sediment – liquefaction

able to use in residential areas and 30 - 40 m vibration
Liquefaction

consequences:boilingofsand

groundproduced by it.

settlement, decrease in soil resistance and strength forkeeping the buildings and bridges up, side ground expansion .

Fig 3 : compaction methods
3-3. Stone Colum method: this method can be used to
Fig 2 : Liquefaction damage

improve soil bearing capacity and decrease the basement267

settlement which is based on the 45 – 35 percent changes ofdecrease and ground bearing strengthening in coastal areas. If

the compaction of unqualified soil by digging wills with

we have used micro pile in loose layer, their ending mustspecific thickness, depth and distance of compaction inreach to the resistant layer.

vertical Colums. Then the materials will be spread in several
layers into the digger will and vibrate and compact by special

According to the received results which are analyses in an

machines. The most significant use of vertical Colums is inarticle by sarang sirafian. About sand soil the following

soft clay that bearing issue appears. While liquefiable soils

conclusions can be achieved.

don’t have this problem.
1. In grouting foundation became of the hitting empty metalframe and also under pressure grout, soil parameters havebeen improved and the weal sand level will change to asuitable and reliable level for bearing of the loaded andaccrued settlement.
2. Considering of some factor such as material expenses,
human force and security beside it, high speed of practicaloperation, sumultaniouy actions of different working groupsof the project and reasonable required machines, using of

Fig 4 : How Stress distribution

micropile foundations in constructing of 15 or less floors

3-4. Soil wills (pressure-breaking wills): the method will
work for liquefiable and soft clay soils. In this procedure atfirst wedding, if underground existing water was problem
well cover the soil after that the inside must be filled withsand, while filling up the cover need to break until thedrainage be done well. After digging for the wills we put
layers on it. The efficiency of this way against earthquake will
be remarkable; its action is easy and can be done in any depth.
Like filter designing method in liquefiable soils that has therole of pressure breaking of water. The between centersshould be 1.5–2 meter. With triangle like models. The
distance is more important rather than the thickness.

would be logical and better economically.
3. High buildings constructing, buildings more than 15 floors
in sand soils due to construction heaviness, micropile
distribution under level, the foundation will become morecompacted and the distances in micropile will decrease andthen individual micropile act like a group micropile finally
this procedure will decrease micropile bearing capacity andwon’t be reasonable.
Conclusions:
1. In some cases soil improvement is the most reasonable
method for solving geotechnical matters.
2. There’s usually or doubt for being liquefiable SPT of 13 –14 and 15, but lower than 12 in any depth soil improvement isrequired.

Fig 5 : With triangle like models
3. For soil improvement of liquefiable soil, methods of
3-5. Grouting method by micropile:

compaction sands will and micropile are being used.

This method should be done before the foundation stage.
Micropile is considered to do two factions, for liquefaction268

4. We use stone Colums and sand wills methods for soft clay
soil.
6. REFERENCES
1. Design and construction ofStone- Colum (2009), dr

mahmood ghazavi,iran .
2. FHWA,( 2000) " micropile design and construction
guidelines", US Department of Transportation, Federal
Highway Administration, NO. - SA - 97 - 070, June,
3. Hausmann, M.R. (1990), “Engineering Principle of
Ground Modifications, Mc-GrawHill Publishing Company,
New York.
4.barksdale, R.D ,and bachus, R.C, design and construction ofstone colum vol FHWA /RD-DESCEMBER-1983

269

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

A Hybrid Intelligent System of Fuzzy C-Mean Clustering and Neural Networks
to Predict the Compaction Characteristics of Fine Grained Soils
Iman Ashayeri, Mahnoosh Biglari
Civil Engineering Dept., School of Engineering, Razi University, Iran
ABSTRACTOne of the most usual processes in geotechnical applications is the compaction of loose fine grained soils to the desireddensity. The results of compaction tests on about 220 plastic soils are collected for this study. The liquid limit andplasticity index of the soils range from 15.5% and 2.1% to 170% and 110% respectively. The aim of this study is topresent a flexible hybrid intelligent system of fuzzy C-mean clustering and neural networks which works with thisdatabase to predict the optimum water content and maximum dry unit weight of soils under standard Proctor effort. The
system does not use a fixed amount of data to train the neural network, but uses the fuzzy C-mean clustering method todetermine the most familiar cases with the unknown data for training the neural network. While, each input question istreated individually a more reliable measure of the performance of the system is obtained.
Keywords: Fuzzy C-mean, Neural Networks, Soil Compaction, Fine Grained Soils

1. INTRODUCTION
In the recent years several investigations were focused on
the applications of intelligent systems in the field of soil
mechanics and geotechnical engineering. Among them theapplications of artificial neural networks on the modelingand prediction of soil behavior or characteristic indices are
numerous. A comprehensive overview on the recent
advances for artificial neural systems in geotechnicalengineering applications can be found in [1]. Most of theseapplications are based on the available experimental or fieldobservations and measurements. The complex, irregularand nonlinear behavior of soil is aimed to be dug out ofavailable data. Although in the most of the applications theperformance of the system was satisfactory, some
important questions still stay and should be considered infuture investigations.
What is the appropriate set of the data to be trained?
How reliable is the system? And how different is thereliability of the system for different inputs?
One of the most usual processes in geotechnicalapplications is the compaction of loose fine grained soils tothe desired density to pass criteria of shear strength,
permeability and etc. The compaction characteristics of finegrained soils are known by the compaction curves fromstandard and modified Proctor tests as the optimum watercontent and the maximum dry unit weight. The results ofcompaction tests on about 220 plastic soils are collectedfrom literature for this study. The aim of this study is topresent a flexible hybrid intelligent system of fuzzy C-meanclustering and neural networks which works with thisdatabase to predict the optimum water content andmaximum dry unit weight of soils under standard Proctor
effort. The system does not use a fixed amount of data to
train the neural network, but uses the fuzzy C-meanclustering method to determine the most familiar cases withthe unknown data for training the neural network. In otherwords, not an identical neural network is trained and used to
predict the optimum water content or maximum dry unit
weight of questioned inputs. While, each input question istreated individually and the most related known patterns areused to answer it, a more reliable measure of theperformance and generalization power of the system isobtained.
2. BACKGROUND OF THE STUDY
Due to the importance of the compaction characteristics offine grained soils in many geotechnical engineering
problems several investigators have attempted to model the
optimum water content and the maximum dry unit weightof the soil at different compaction efforts. Among them[2]-[7] can be categorized in traditional mathematicalmethods used to predict optimum water content and/or
maximum dry unit weight in terms of index properties andcompaction effort. In addition to the traditional methods
some investigators utilized artificial neural networks topredict the compaction characteristics [8]-[12]. Although
the approaches differ in each study they resulted in
acceptable performance of the system within the range ofthe data used for correlating the empirical equations or
training the artificial neural networks.
Furthermore, Shahin et al. [13] proposed using fuzzy
C-mean clustering in selection of sets of data to train,validate and test artificial neural networks. Fuzzy C-mean(FCM) clustering uses fuzzy logic in clustering data so thateach data pattern belongs to each cluster with a membershipvalue. The method was first developed by Dunn [14] andimproved by Bezdek [15]. The FCM clustering gives backthe membership value of each data pattern to each cluster aswell as the centers for the clusters.
3. CHARACTERISTICS OF DATABASE
The database required for this study was collected fromhigh quality research articles previously published byseveral investigators across the globe. Table 1 presents the270range of index properties of the soils available in thisdatabase.
Table 1. Range of index properties of soils in database
Index
CSF (%)FC (%)Gs
LL (%)
Min
2
18.6
2.46
15.5
Max
98
100
2.90
170
CSF: clay size fraction, FC: fine contents

PI (%)
2.1
110

Fig. 1 plots the plasticity index of the soils versus theirliquid limits in the Casagrande chart. Furthermore, thevalues of optimum water content (OWC), and maximum
dry unit weight (MDUW), of the soils at standard Proctor(SP) compaction effort are plotted against liquid limit as inFigs. 2 and 3 respectively. It is clear from the Table 1 andFigs. 1 to 3 that the variety of data range plus the inherent
complex influence of the indices on OWC and MDUW ofsoils will make it inaccurate to postulate a unique model for
prediction of OWC and MDUW, which works in the whole
data range in the database. One may ask the first question
mentioned in the introduction of what the most similar data
patterns are to specific unknown? Furthermore, theaccuracy and reliability of a unique model trained for alldata range in the database is variable with respect to the
input record. Therefore, one concludes that if only the mostsimilar patterns to the unknown record are trained to themodel then the answer of the trained model would be morereliable, which addresses the second question mentioned in
the introduction. These two main concepts are dealt withthe selection of most similar patterns to the input question
using FCM. The performance of this hybrid application of
FCM and ANN is validated by two question records in thefollowing sections.

Fig. 1 LL and PI of soils in the database

4. VALIDATION OF THE PERFORMANCE OF FCM
AND ANN COMBINATION
Two question records of low plastic (LPS), and high plastic(HPS), soils are selected to validate the application of thehybrid FCM and ANN model. Table 2 summarizes theindex properties as well as OWC and MDUW of them.
Table 2. Index and compaction characteristics of validation set
Index Silt
CSF
Gs
LL
PI
OWC MDUW
(%)
(%)
(%)
(%)
(%)
(KN/m3)
LPS
60
18
2.65
29
12
15.4
17.4
HPS
45
25
2.49
52
31
19.3
16.6

Fig. 2 OWC vs. LL of soils in database

Fig. 2 MDUW vs. LL of soils in database

To seek simplicity and avoid inclusion of the effects of the
ANN architecture on their performance, the architecture of
the ANNs are kept similar that is a single hidden layer
feedforward error backpropagation network with 6 input
neurons corresponding to the 6 index parameters of Silt
content, CSF, FC, Gs, LL and PI, with two neurons in the
only intermediate layer and finally one neuron for the
output, which is OWC or MDUW.
With respect to the contribution of A. Casagrande on theclassification of plastic soils based of their LL and PI, thesetwo variables are used to classify the patterns in the
database by FCM. Each question record is classified withthe whole database several times considering the number ofclusters is variable from 3 to 6. A computer program was
prepared in the MatLab environment to give the training set
of ANN corresponding to each question record. The ANN
is trained and the performance of the training phase isevaluated by the average of training absolute error (ATErr),
as well as the standard deviation of the training absolute
error (SDTErr). Afterward, the question record is recalled
and the predicted value by the model is compared by theactual output quantity given as in Table 2 by calculating theabsolute error. This process is repeated for each question
record at the other number of clusters. For instance, Fig. 4
presents the selected training set for the LPS when thenumber of clusters is 5 and Fig. 5 presents the selectedtraining set for HPS when the number of clusters is 4 in the
Casagrande chart with the position of the LPS and HPS as
well as the centers of the clusters. Tables 3 and 4
summarize the statistical quantities identified above for
LPS and HPS at all combination of cases respectively.

271

5. CONCLUDING REMARKSThe performance of the ANN models at training phase isevaluated by the ATErr and SDTErr values so that thesmaller values are more preferred. It can be found from theTables 3 and 4 that the smaller values of ATErr and SDTErr
correspond to the cases with the largest MMV. Looking to
the number of training patterns reveals that in all cases thishigher performance level for the ANN models wasachieved by smaller size of the training set. In other words,
the corresponding training set with smaller ATErr andSDTErr contains the most correlated patterns in the
database required to train the ANN.
One measure for the reliability of the performance of the
ANN is the smaller error at testing set. However, this is notknown when actually an unknown record is answered by
the model. This shortcoming can be fulfilled by a quantity
which expresses the correlation between the training set andthe unknown record in terms of their known input
parameters. In this study this correlation is expressed by the
MMV. Comparing the MMV and Err of each ANN modelfor question records in Tables 3 and 4 reveal that thesmaller error in predicted values of OWC and MDUW also
corresponds to the largest MMV. It can be concluded thatthe combination of the statistical measures from training
phase with the quantity obtained from FCM analysis canpresent a suitable criterion of the performance andreliability of the ANN. It should be noted that analysis
performed in this study reveals that increasing the numberof training patterns does not increase the performance andreliability of the system. However, the more appropriate
and coherent training set with the question record results inmore reliable answers and this study shows that such
coherency and appropriateness can be expressed anddetermined by Fuzzy C-mean clustering technique.

Fig. 4 Selected training set for LPS with 5 clusters

Fig. 5 Selected training set for HPS with 4 clusters

OWC

Table 3. Summary of statistical quantities of models for LPS
Model
No.of
No.of
ATErr SDTErr MMV
Errclusters training
(%)
(%)
of
LPS
patterns
LPS
(%)
3
29
9.6
8.1
0.97
6.8
4
25
9.5
8.3
0.97
7.2
5
12
6.4
3.6
0.99
1.8
6
24
9.3
7.3
0.79
4.2
3
30
3.8
2.4
0.97
2.7
4
24
2.3
1.7
0.975
2.2
5
12
1.7
0.9
0.98
0.2
6
23
1.8
1.2
0.74
0.5

6. ACKNOWLEDGMENT

MDUW

The authors wish to acknowledge the support of Razi
University by funding this research project.
7. REFERENCES

MDUW

OWC

Table 4. Summary of statistical quantities of models for HPS
Model
No.of
No.of
ATErr SDTErr MMV
Errclusters training
(%)
(%)
of
LPS
patterns
LPS
(%)
3
17
10.0
7.0
0.84
26.3
4
15
5.8
3.9
0.98
2.5
5
19
6.5
4.9
0.63
17.2
3
45
3.9
2.4
0.65
4.1
4
15
1.7
1.8
0.98
3.5
5
19
2.7
2.2
0.59
6.0
6
18
2.6
2.0
0.80
7.4

In addition to statistical quantities for training and testing of
the neural networks the maximum membership value(MMV), of the question records (LPS or HPS) are given in
Tables 3 and 4 for each case. As the number of the clusters
increases the higher MMV indicates that the question
record is closer to the center of the cluster. The analysis on
the results is presented in the following section.

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Metropolitan Area” J. Soil Mech. And Found. Div., vol.
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[3] Hilf J, “A Rapid Method of Construction Control for
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vol. 26, 1956.
[4] Ring G, Sallberg J, and Collins W, “Correlation ofCompaction and Classification Test Data” Hwy. Res.
Bull. 1962, pp. 55-75.
[5] Ramiah B, Viswanath V, and Krishnamurthy H,“Interrelationship of Compaction and Index
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272

[6] Wang M, and Huang C, “Soil Compaction andPermeability Prediction Models” J. Envir. Engrg., vol.
110, 1984, pp. 1063-1083.
[7] Blotz LR, Benson CH, and Boutwell GP, “Estimating
Optimum Water Content and Maximum dry Unit
Weight for Compacted Clays” J. Geotech. Geoenvir.
Engrg., vol. 124, 1998, pp. 907-912.
[8] Basheer IA, “Empirical Modeling of the Compaction
Curve of Cohesive Soils” Can. Geotech. J., vol. 38,
2001, pp. 29-45.
[9] Yasrebi SS, and Ashayeri I, “Predicting Optimum
Water Content and Dry Unit Weight of Clay Soils
Using Neural Networks” in Proc. 56th Canadian
Geotechnical Conference, 2003.
[10] Sinha SK, and Wang MC, “Artificial Neural Network
Prediction Models for Soil Compaction andPermeability”, Geotechnical and GeologicalEngineering, vol. 26, 2008, pp. 47-64.
[11] Adel-Rahman AH, “Predicting Compaction of
Cohesionless Soils Using ANN”, Ground
Improvement, vol. 161, 2008, pp. 3-8.
[12] Sivrikaya O, Soycan TY, “Estimation of Compaction
Parameters of Fine-grained Soils in terms ofCompaction Energy Using Artificial Neural
Networks”, Int. J. Numer. Anal. Meth. Geomech., vol.
35, Dec. 2011, pp. 1830-1841.
[13] Shahin MA, Maier HR, and Jaksa MB, “Data Division
for Developing Neural Networks Applied togeotechnical Engineering”, J. Computing in Civil
Engrg, vol. 18, 2004, pp. 105-114.
[14] Dunn JC, “A Fuzzy relative of the ISODATA Process
and Its Use in Detecting Compact Well-separated
Clusters”, J. of Cybernatics, vol. 3, 1973, pp. 32-57.
[15] Bezdek JC, “Pattern Recognition with Fuzzy Objective
Function Algotrithms”, Plenum Press, 1981.

273

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Equivalent Linear Seismic Ground Response Analysis of
Unsaturated Soil Deposits
Mahnoosh Biglari, Iman Ashayeri
Civil Engineering Dept., School of Engineering, Razi University, Iran
ABSTRACT
Seismic ground motion is profoundly affected by geometrical and mechanical properties of soil deposits overlaying
bedrock. Seismic ground response of saturated soil deposits was studied in literature by applying the effects of soil
stress state and index properties on the strain-dependent normalized shear modulus reduction and damping ratio curves
in an equivalent linear analysis. Recent experimental investigations revealed that, G 0 , G/G 0 , and D of unsaturated soils
are influenced by stress state as well as suction. This study presents the results of linear and equivalent linear seismic
ground response analysis of unsaturated soil deposits with considering suction effects on G/G 0 and D curves. Analysis
was done by the computer program EERA for two sets of soil profiles, which include in saturated, constant and linearlyvariable suction deposits. Results show that natural frequency of the soil profile increases as the suction increases.
Keywords: soil, soil dynamics, equivalent-linear, seismic ground response analysis

1. INTRODUCTION
The local soil conditions have a profound influence on the
ground response during earthquakes, which is modelled
throughdirectnon-linear
elasto-plastic[1]orequivalent-linear elastic ground response analysis [2] and [3].
In spite of its theoretical shortcomings, the latter has become
the major tool in practical engineering applications due to its
simplicity. The equivalent-linear analysis consists ofmodifying normalized shear modulus G/G 0 , and damping
ratio D, of a visco-elastic soil model with shear strain level, γ.
Regularly, the strain-dependent normalized shear modulus
reduction (G/G 0 -γ), damping ratio (D-γ) curves are obtainedby laboratory tests. However, mathematical functions were
presented for (G/G 0 -γ) and (D-γ) in terms of stress state andindex properties of saturated soils based on enormous
experiments by various investigators [4]-[6].
Experimental investigations revealed that, the initial shear
modulus G 0 , of unsaturated soils is influenced by stress state
as well as suction [7]-[12]. Furthermore, recent investigationon the measurements of shear modulus of an unsaturated soil
at wider shear strain range by suction-controlled cyclic
tri-axial apparatus show that G-γ and D-γ curves are
influenced by the suction levels too [13]. On the basis of thisexperimental evidence, Biglari and Ashayeri [14] modified
the empirical equation proposed by [6] for G/G 0 -γ and D-γ totake into the account the influence of the suction level as well
as stress state and index properties for unsaturated soils.
d’Onza et al. [15] investigated the effects of unsaturated soil
suction on the seismic ground response by means of linear
analysis which takes into the account the influence of suction
on the shear wave velocity or initial shear modulus of theunsaturated soil deposits. They have presented the results interms of amplification ratio of surface motion in thefrequency domain and concluded that the natural frequency
of soil deposit significantly increases with suction increase
and the maximum amplification ratio is substantially reduced.
This study presents the results of linear and one dimensional
equivalent linear seismic ground response analysis of
unsaturated soil deposit not only considering the effects ofsuction variation on the shear wave velocity (or G 0 ), but also
taking into the account the dependency of G/G 0 -γ and D-γ on
the suction level based on the developed empirical model forG-γ and D-γ [14]. Seismic ground response analysis was donewith the computer program “Equiva-lent-linear Earthquakesite Response Analyses” (EERA) [3], for five soil suction
profiles and three time histories of acceleration. The results
are discussed in terms of comparison between theacceleration time history at surface and bedrock, theamplification ratio and natural frequency of soil deposits aswell as spectral acceleration of the ground motion at surface
and bedrock.
2. SOIL PROFILESFive soil profiles were considered in this study within two
sets of analysis. The first set compares two soil profiles P1
and P2 of lean clay deposit 24 meters deep overlaying the
bedrock. Table 1 presents the variation of suction, shear wave
velocity, and total unit weight of different layers of the soil
profiles P1 and P2. Soil profile P1 is formed of a two metersof saturated soil layer over the bedrock that is overlaid by 22
meters of unsaturated soil with variable suctions. The
unsaturated soil deposit of P1 was divided into 6 layers ofconstant suction varying from zero to 250 kPa. On the other
hand, soil profile P2 is formed of 24 meters of the saturated
soil deposit overlaying the bedrock that is divided into 7
layers of variable shear wave velocity and unit weight.
The second set of analysis compares three soil profiles P3, P4,274

and P5 of the same deposit 24 meters deep overlaying the
bedrock. Table 2 presents the variation of suction, shear wave
velocity, and total unit weight of the different layers of thesoil profiles P3, P4, and P5. Soil profile P3 is formed of 24
meters of the saturated soil deposit overlaying the bedrock
that is divided into 6 layers with variable shear wave velocity
and unit weight. Soil profile P4 is formed of 24 meters of theunsaturated soil deposit with constant suction of 150 kPa
overlaying the bedrock and is divided into 6 layers withvariable shear wave velocity and unit weight. Soil profile P5
is formed of 24 meters of the unsaturated soil deposit withconstant suction of 300 kPa overlaying the bedrock and again
is divided into 6 layers with variable shear wave velocity andunit weight.

Fig. 1 G/G 0 -γ and D-γ curves of layers of P1

Thickness
(m)

Layer

Table 1. Description of soil profiles of the first set of analysis
Profile
P1
P2
s1

Vs2

γ3

Vs

γ

(m/s)

(KN/m3)

1
4
250
159
13.43
0
2
4
200
159
13.63
0
3
4
150
170
14.18
0
4
4
100
187
15.15
0
5
4
50
201
16.19
0
6
2
0
195
19.86
0
7
2
0
200
19.95
0
8
bedrock
1219
22.02
1
Suction, 2Shear wave velosity, 3Unit weight

125
125
130
142
154
164
172
1219

19.30
19.30
19.38
19.57
19.61
19.65
19.77
22.02

(m/s)

(KN/m3)

s

(kPa)

(kPa)

Fig. 2 G/G 0 -γ and D-γ curves of layers of P2

Vs
(m/s)

γ
(KN/m3)

Vs
(m/s)

γ
(KN/m3)

4
4
4
4
4
4
bed

γ
(KN/m3)Thickness (m)

1
2
3
4
5
6
7

Vs
(m/s)

Layer

Table 2. Description of soil profiles of the second set of analysis
ProfileP3 (s=0)P4 (s=150)
P5 (s=300)

125
125
130
142
154
168
1219

19.30
19.30
19.38
19.57
19.61
19.71
22.02

159
159
170
188
203
219
1219

13.70
13.70
14.20
14.60
15.10
15.30
22.02

176
176
185
203
219
233
1219

13.50
13.50
13.80
14.30
14.50
14.70
22.02

The shear wave velocities of the soil layers of all the fiveprofiles were calculated from the model for initial shear
modulus of unsaturated soils calibrated for the current
material, which was presented by [11] and [12].
3. MODULUS REDUCTION AND DAMPING CURVES
In order to generate the modulus reduction and damping
curves, the empirical models presented by [14] were usedwith considering the soil plasticity index (PI = 12%), andappropriate suction and confining pressure of each layer ofthe soil profiles. Figs. 1 to 5 present the generated G/G 0 -γ andD-γ for layers of soil profiles P1 to P5, respectively.

Fig. 3 G/G 0 -γ and D-γ curves of layers of P3

4. GROUND RESPONSE ANALYSIS
Seismic ground response analysis was done with thecomputer program EERA [3] for the five soil profiles P1 to P5.
EERA is a modern implementation of the well-known
concepts of equivalent linear earthquake site response
analysis applied in SHAKE [16]. The soil profiles are
subjected to the input ground motion from the bedrock that is
specified as an out-crop motion. To avoid dependency of theground response to the input motion, three earthquakes were275

Amplification Ratio

Table 3. Comparison between PGA values of bed rock and surfaceof the all profiles
PGA (g), Amplification factor, 1 st natural frequency (Hz)ProfileLoma Prieta
Kobe
ChiChiP1
0.146, 2.25, 1.4
P2
0.113, 2.29, 0.8
P3
0.113, 2.29, 0.8
0.117, 2.30, 0.8
0.104, 2.19, 0.8
P40.148, 2.61, 1.4
0.155, 2.56, 1.4
0.154, 2.59, 1.4
P50.146, 2.71, 1.6
0.154, 2.69, 1.6
0.175, 2.69, 1.6

3
1.4 HzP1
P2

2
0.8 Hz
1
0
0

5

10
15
Frequency (Hz)20

25

Fig. 6 Amplification functions of P1 and P2.

Amplification Ratio

used in the analysis. These three input motions are theacceleration time history of Loma Prieta 1989, Kobe 1995,
and Chichi 1999 earth-quakes (Records P0782, P1043 &P1116 at http://peer.berkeley.edu/smcat/). The groundmotions are normalized to a target peak acceleration of 0.1g.
The comparisons between Peak Ground Acceleration (PGA)at surface, amplification factor, and first natural frequency ofeach soil profile are presented in Table 3 for the three groundmotions. The first natural frequency increases from 0.8 Hz for
P2 to 1.4 Hz for P1 and it varies from 0.8 Hz for P3 to 1.6 Hz
for P5 by increasing suction. The PGA of the motion at
surface of the all profiles is larger than the bedrock motion
and this is more significant at the unsaturated profiles.
Figure 6 shows amplification between the surface motion andthe base motion at varying frequencies for P1 and P2 forLoma Prieta earthquake. The increase in the natural
frequency with suction increase reasonably can be assigned tothe in-crease of shear modulus by suction. However, thevariation of amplification factor is related to the admittanceratio between the soil and the bedrock, which resulted into
increasing amplification factor by increasing suction. To
provide comparison of the effects of non-linearity, series oflinear analyses were performed for P3 to P5 which, the results
are presented in figures 7 to 9. It is clear that non-linearity has
influenced the results significantly in terms of both
amplification factor and natural frequency.

6
Equivalent Linear Analysis

1.6 Hz5
4Linear Analysis

0.8 Hz

3
2
1
0
0

5

10

15

20

25

Frequency (Hz)Fig. 7 Amplification functions of P3 in different analysis.

Amplification Ratio

6
Fig. 4 G/G 0 -γ and D-γ curves of layers of P45

2.0 Hz4

1.4 HzEquivalent Linear Analysis
Linear Analysis

3
2
1
0
0

5

10

15

20

25

Frequency (Hz)Fig. 8 Amplification functions of P4 in different analysis.

Fig. 5 G/G 0 -γ and D-γ curves of layers of P5

Another index to seismic hazard of buildings and structures is
response spectral acceleration curve. Spectral acceleration S a ,is the maximum acceleration of a single degree of freedom
oscillator with different natural frequencies but uniquedamping. Figures 10 to 13 compare the S a with 5% damping
276

Spectral Acceleration (g)ratio at the ground surface of the profiles with the one at
bedrock for the two sets of analysis.
Shift of S a to the lower frequencies (higher periods) at thesurface of the saturated profiles (P2 and P3) is observed.
However, S a of the surface ground motion of unsaturated
profiles (P1, P4, and P5) is considerably different with theones of saturated profiles (P2 and P3) and the bed rock. The
results show that the S a increases by suction increase for highfrequencies (periods less than 1 sec). This is very importantand should be noticed in seismic design of structures on theun-saturated soil deposits particularly for short buildings withhigh natural frequency.

0.7
0.6
0.5
0.4

Bed Rock
P3
P4
P5

0.3
0.2
0.1
0
0.01

0.1

1

10

Period (sec)6
2.2 Hz5

Equivalent Linear Analysis

0.7

Linear Analysis

4Spectral Acceleration (g)Amplification Ratio

Fig. 12 S a of P3, P4, and P5 for Kobe.

1.6 Hz

3
2
1
0
0

5

10

15

20

25

Frequency (Hz)Fig. 9 Amplification functions of P5 in different analysis.

0.6

Bed Rock
P3

0.5

P4

0.4

P5

0.3
0.2
0.1
0
0.01

0.1

1

10

Period (sec)Spectral Acceleration (g)Fig. 13 S a of P3, P4, and P5 for Chichi.

0.7
0.6
0.5

5. SUMMARY AND CONCLUSIONS

Bed Rock
P1
P2

0.4
0.3
0.2
0.1
0
0.01

0.1
1
Period (sec)10

Spectral Acceleration (g)Fig. 10 S a of P1 and P2 for Loma Prieta.
0.7
0.6
0.5
0.4

Bed Rock
P3
P4
P5

0.3
0.2
0.1
0
0.01

0.1

1
Period (sec)Fig. 11 S a of P3, P4, and P5 for Loma Prieta.

10

The aim of this study was to perform 1D equivalent linear site
response analysis by considering the influence of suction on
the shear wave velocity, shear modulus reduction anddamping ratio curves. Two sets of analysis compare theresponse of three unsaturated soil profiles with the saturated
ones due to three earthquakes. It came out with the following
conclusions for the selected soil profiles.
• The natural frequency of the soil profile increases as the
suction increases. This will attract the attentions to thesuitable natural frequency of the buildings on the unsaturated
soil deposits.
• The ground motions were amplified at the surface for allof the saturated (P2 and P3) and unsaturated profiles (P1, P4& P5). However, the combined conditions considered in thesoil profiles showed that the amplification ratio is slightly
larger at unsaturated soils that may not to be a generalbehaviour.
• Response spectral accelerations calculated for dampingof 5%, show that the response of the structures on theunsaturated soil profiles are more severe than saturated ones,
particularly at higher frequencies.
6. ACKNOWLEDGMENT
The authors wish to acknowledge the support of the Razi
University through research grant Code: 36471.

277

7. REFERENCES[1] Hashash Y M A, and Park D, “Non-linear
one-dimensional seismic ground motion propagation in
the Mississippi embayment”, Engineering Geology, vol.
62, 2001, pp.185-206.
[2] Idriss I M, and Seed H B, “Seismic Response of
Horizontal Soil Layers”, J. Soil Mech. Found. Div., vol.
94, 1968, pp.1003-1031.
[3] Bardet JP, Ichii K, and Lin CH, “EERA, A computer
program for Equivalent linear Earth-quake site Response
Analysis of layered soils deposits”, Software Manual,
University of Southern California, Los Angeles, 2000.
[4] Seed HB, Wong RT, Idriss IM, and Tokimatsu K“Moduli and damping factors for dynamic analyses ofcohesionless soils”, J. Geotech. Engrg., vol. 112, 1986,
pp.1016-1032.
[5] Vucetic M, and Dobry R, “Effect of soil plasticity on
cyclic response”, J. Geotech. Engrg., vol. 117, 1991,
pp.89-107.
[6] Ishibashi I, and Zhang X, “Unified dynamic shear moduli
and damping ratios of sands and clays”, Soils andFoundations, vol. 33, 1993, pp.182-191.
[7] Mancuso C, Vassallo R, and d’Onofrio A, “Small strain
behavior of a silty sand in controlled-suction resonant
column – torsional shear tests”, Can. Geotech. J., vol. 39,
2002, pp.22– 31.
[8] Vassallo R, Mancuso C, and Vinale F, “Effects of net
stress and suction history on the small strain stiffness of acompacted clayey silt”, Can. Geotech. J., vol. 44, 2007,
pp. 447- 462.
[9] Vassallo R, Mancuso C, and Vinale F, “Modelling theinfluence of stress–strain history on the initial shear
stiffness of an unsaturated compacted silt”, Can. GeotechJ., vol. 44, 2007, pp.463- 472.
[10] Ng CWW, and Yung SY, “Determination of theanisotropic shear stiffness of an unsaturated decomposedsoil”, Géotechnique, vol. 58, 2008, pp.23–35.
[11] Biglari M, d’Onofrio A, Claudio M, Jafari MK, andShafiee A, “Small strain stiffness of Zenoz kaolin in
unsaturated conditions”, Can. Geotech. J., vol. 49, 2012,
pp.311-322.
[12] Biglari M, Mancuso C, d’Onofrio A, Jafari MK, andShafiee A, “Modelling the initial shear stiffness of
unsaturated soils as a function of the coupled effects ofthe void ratio and the degree of saturation”, Computers
and Geotechnics, vol. 36, 2011, pp.709-720.
[13] Biglari M, Jafari MK, Shafiee A, Mancuso C, andd’Onofrio A, “Shear Modulus and Damping Ratio of
Unsaturated Kaolin Measured by New
Suction-Controlled Cyclic Triaxial Device”,
Geotechnical Testing J., vol. 34, 2011, pp.525-536.
[14] Biglari M, and Ashayeri I, “An empirical model for shear
modulus and damping ratio of unsaturated soils”, in Proc.
5th Asia-Pacific Conf. on Unsaturated Soils, 2011,
pp.591-596.
[15] d’Onza F, d’Onofrio A, and Mancuso C, “Effects of
unsaturated soil state on the local seismic response of soil
deposits”, in Proc. 1st European Conf. on Unsaturated
Soils, 2008, pp.531-536.

[16] Schnabel PB, Lysmer J, and Seed HB, “SHAKE: A
Computer Program for Earthquake Response Analysis of
Horizontally Layered Sites”, Report No.
UCB/EERC-72/12, Earthquake Engineering ResearchCenter, University of California, Berkeley, December
1972, 102p.

278

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The Correction of Shallow Foundation Settlement Equations
by Finite Element Method
Mohammad Arab Ameri1, Hakime Rabbanifar2 and Mehdi Arab Ameri3
1MSc, earthquake engineering, Iran2
MSc, geotechnique engineering, Iran
3
MSc, geotechnique engineering, Iran
P

P

P

P

P

P

P

P

P

P

P

ABSTRACT
Since Hough (1959) showed that foundation settlement calculations must done after foundation bearing capacity
calculations, several methods have been suggested to determine settlement of shallow foundations. These methods
have several limitations such as hypothesis of vertical strain creation due to foundation loading, one-dimensional(vertical) excess pore water pressure dissipation and irrespective of horizontal drainage of the pore water. Also thesemethods do not account for the effect of soil type on amount of excess pore water pressure beneath the footing andeffect of plastic deformations in soil structure. In this study the effective elements in settlement of shallow foundations
have been investigated and with the aid of PLAXIS software, the effect of elastic strains, plastic strain related to local
collapse, anisotropy in soil deformation parameters and secondary strains have been considered in modeling thesettlement of shallow foundations with footing dimension and soils with different overconsolidation ratio. Then the
results have been introduced as modification coefficients.
Keywords: Settlement, Shallow Foundations, Correction Coefficient, Relative Pressure, Secondary Consolidation

1. INTRODUCTION
Nowadays the fact that the settlement of shallow foundations
is an important factor in their design is widely accepted. Thisfactor earned its situation in design of these foundations notearlier than 1960. Prior practical engineers thought that if afoundation is approved against the bearing capacity issue itwill has enough safety factor against the settlement issue. It
was 1959 when Hough indicated that the estimation of
settlement in shallow foundations should be followed rightafter the consideration of bearing capacity of suchfoundations. Since then, several methods for this estimation
are introduced.
The first point of view for estimation of settlement in shallow
foundations is related to Terzaghi (1925) who adopted theparameters of his model from one-dimensional oedometer
test. Despite its widespread usage, the method worked withsome limitations. In this theory, it was assumed that: The
foundation pressure produces only vertical strains in subsoil,
all the induced settlement is one- dimensional and that
relating to the shear strain is neglected, the dissipation ofexcess pore water pressure takes place only in verticaldirection and the lateral dissipation is ignored and finally theexcess pore water pressure is equal for all types of subsoil andis not dependant to the soil type.
Above-mentioned limitations of Terzaghi method issomewhat revised at the method introduced by Skempton and
Bjerrum (1957). They presented the settlement of a shallow
foundation as a resultant of two phenomena: Immidiate andconsolidation settlement along with a correction coefficientwhich is known as Skempton and Bjerrum coefficient.
Similar to the Terzaghi method, their work have several
limitations, for instance: Skempton and Bjerrum limited theirstudies on strip and circular foundations and other shape offoundations like rectangular foundations were not put into
consideration, they assumed no specific distinction between
rigid and flexible foundations and finally the plastic
deformations of soil mass was neglected and all thesettlement assumed to be elastic where as in coarse-grainedsoils, the secondary plastic settlement may become several
times greater than elastic one. Regarding these limitations,introducing a new method with revisions on these limitations
seems to be unavoidable.
Studies of Davis and Poulos (1968) indicated that foroverconsolidated and soft clays, one-dimensional
consolidation analysis represents a good approximation for
total settlement and this approximation is valid for soils withdrained poisson's ratio less than 0.35. Increasing the value of
this parameter and sickness of the subsoil strata, make themethod to under predict the actual settlement value.
Most of the commonly-used methods of settlement analysis,
assume the soil to behave elastically. Since the nonlinearity insoil behavior is considered in other ones and normally andoverconsolidation states is distinguishable in them by
employing different values of compressibility for these twostates, the consideration of local yielding arising from thefoundation pressure is not noticeable in these methods(Poulos, 2000). Burland et al. (1977) showed that fornormally consolidated soils neglecting the local yielding andplastic deformation phenomena make the settlement analysis
to be unreliable in most of the cases.
Davis and Poulos (1968) indicated that applied pressure atwhich local yielding commences is a function of safety factor
against the bearing capacity issue and initial stress state.
D'Appolonia (1971) extended this idea and presented acorrection factor for consideration the influence of local
yielding. It is important to notice that the effect of local
yielding on consolidation is not put into consider since the279studies of Carter et al. (1979) show that local yielding has no
meaningful effect on consolidation settlement and elastic
theory introduces appropriate approximation in this issue.
The existence of creep and secondary consolidation, havestrict influence on the rate and magnitude of settlement offoundations on clay soil. Most of the practical methods which
regard the creep in the settlement analysis are related tostudies done by Buisman (1936). In these studies, a linear
relation between the magnitude of settlement and thelogarithm of time is assumed. The gradient of thisrelationship is known as the coefficient of secondary
consolidation, cα.

cα = ∆E∆(Logt )(1)

Mesri and Godlewski (1977) found that cα is nearly
dependant to the consolidation coefficient, cc and the ratio ofcα/cc has a narrow range of variation. The rate of secondary
consolidation for an overconsolidated soil is much less than
the normally consolidation state of that soil.
The difficulty of employing cc in settlement analysis arises
from this fact that the time at which the creep commences is
not well definable. Since some researchers believe that creep
takes place after the consolidation (Mesri et al. 1994) others
assume a simultaneous occurrence of creep and initialconsolidation (Leroueil, 1996).
Since different values of secondary consolidation coefficientsare seen in the literature (Gibson and Lo 1961, Garlanger1972), employing appropriate parameters for including thecreep in the settlement analysis at least in its simplest and
one-dimensional model are difficult and unusual. It seems
that, among the different aspects of settlement theory, creep
and secondary consolidation have the least development andneed for extra considerations.
The above mentioned considerations approve that, in the new
methods for settlement analysis, the influence of both elastic
and plastic strains should be considered, where the hardening
soil model which is employed in the PLAXIS software
competently has this ability.

compression hardening. Shear hardening and Compression
hardening are used to model irreversible strains due torespectively primary deviatoric loading and primary
compression in oedometer and isotropic loadings.
This model is an advanced model for simulating the
behaviour of different types of soil, both soft and stiff soils,(Schanz, 1998).
When a soil is subjected to initial deviatoric stress, indicates asimultaneous decrease in its hardness and occurrence ofirreversible plastic strain. For example, for drained triaxial
test, the relation between axial strain and deviatoric stress canbe described by:− ε1 =

1q
Ei 1 − q q a(2)
Where ε1 is vertical strain, q is the deviatoric stress; qa is theasymptotic value of the shear strength and Ei the initialstiffness.
Such relationship first formulated by Konder (1963) and is
employed at hyperbolic model (Duncan and Chang, 1970).
The Hardening Soil model, however, supersedes thehyperbolic one firstly by using the theory of plasticity rather
than the elastic one. Secondly by including soil dilatancy andthirdly by introducing a yield cap.
Some of the specifications of this model are: the dependenceof hardness to the stress with a power law, the dependence ofplastic strains to initial deviatoric loading and primary
compression, elastic unloading / reloading and finally failure
according to the Mohr-Coulomb model.
The basic feature of this model is its dependency on the stresslevel. For instance, for oedometric stress and strain condition
the model introduces an equation as:(

2. Present Study
Nowadays the finite element theory and softwares using it
have an important part in the analysis of stress and strain
inside a soil mass. The output parameters of such softwares
have a good approximation with the monitored parameters.
At present study, with the aid of PLAXIS software andemploying Hardening Soil model, the settlement of shallow
foundations which are assumed to be rigid are estimated andthe results are compared with those obtained from elastic
solutions and sets of correction factors are introduced forelastic settlement analysis. In this study, the modeled
foundations have three different widths and are embedded onfour clays with different overconsolidation ratios.
The Hardening Soil Model has no constant yield surface andcan expand due to plastic strain. Two types of hardening are
distinguishable in this model, namely shear hardening andrefEoed = Eoed
− σ '1 / p ref

)

m

(3)
Where Eoed is oedometer modulus, pref is a reference

E refpressure, oed is oedometer modulus measured at reference
pressure, σ'1 is the major principal stress and m the power instress-dependent stiffness relation.
3. The Employed Model

The geometry of the model used in this study ispresented in fig. 1. The aspect ratios are assumed to beconstant as H/B= 5 and L/B= 9. For considering theinfluence of the foundation width in the analysis, threevalues of 1, 1.5 and 2 are assumed for this parameter andsince the effect of foundation flexibility is neglected andis relegated to future studies, a high value of bending280

modulus is assumed for the foundation to demonstrate arigid behavior.
B

L
Fig. 1- Schematic shape of current model

In current study, four types of clays with different
overconsolidation ratios are considered with parametersdescribed in table I.
Table I. The parameters of the employed soils
Soil
NC OCR=1.7 OCR=2.5 OCR=4
Parameters
3.2
1.6 E 43.75 E 4 7.5 E 4E50refE3
3.2
ref
1.6 E 4
3E4
6E4
Eoed
E3
9.6
2.25 E
4.8 E 49E4Eurref
E3
5
ν
0.3
0.3
0.3
0.3
m
0.8
0.65
0.6
0.55
C
1
7
11
20
φ
24
25
25
25
ψ
0
0
0
0
γ unsat
16
16
16
16
γ sat
18
18
18
18
1.0
1.0 E kE1.0 E - 4
1.0 E - 4
4
4
k0
0.59
0.75
0.91
1.15
R

R

E50ref (kN/m2) is Secant stiffness in standardrefEoed(kN/m2) is tangent stiffness for
P

P

E ref

ur (kN/m2) is Unloading
primary oedometer loading,/reloading stiffness, ν is Poisson's ratio, m is Power for
stress-level dependency of stiffness, c (kN/m2) is effective
cohesion, φ is effective angel of internal friction, ψ is theangel of dilatancy, γunsat (kN/m3) is the unsaturated
volumetric weight, γsat (kN/m3) is the saturated volumetricweight, k is the bulk modulus and k0 coefficient of lateralearth pressure.
P

P

P

P

P

P

(4)
Where Si, Sc and Ss are the immediate, consolidationand secondary consolidation settlements respectively
and μ and η are the correction coefficients for theimmediate and consolidation settlements respectively.
All the settlement components are calculated using the
PLAXIS software.
The correction factors for the immediate andconsolidation settlements are composed of two
components: the correction factor of using foundation
width less than 2m, I and the correction factor of
employing elastic theory in the analysis, β.

µ = Ii βi,η = Ic βc(5)Where:
Ii= the correction factor for using elastic theory in theimmediate settlement analysis
βi= the correction factor for using foundation width lessthan 2m in the immediate settlement analysis
Ic= the correction factor for using elastic theory in the
consolidation settlement analysis
βc= the correction factor for using foundation width lessthan 2m in the consolidation settlement analysis
In the current analysis, foundation loading is continued
up to the subsoil failure. It is evident that the pressure atwhich the failure occurs, in other words the bearing
capacity of the foundation is related to foundation
aspects. For omitting the effect of this issue, the underpressure in the foundation is divided by its bearing
capacity as:(6) RP =R

drained triaxial test,

The general equation for settlement analysis is
introduced as:

S = µ Si +η Sc + S s

HIn table I, the4. Results

P
BC

Where RP is the relative pressure and according to its
definition is the negative power of safety factor against
the bearing capacity of the foundation, P is the applied
load and BC the bearing capacity of the foundation. For
defining the correction factor of using elastic theory in
the immediate settlement analysis, I i the result ofmodeling where the Hardening Soil model is used isdivided by those whom the elastic theory is adopted.
The results are smoothened and introduced as the I i in
Fig 2. For other overconsolidation ratios, theextrapolation should be employed.

P

P

281

R

R

R

R

0.9
NC

4

OCR=1.7

3.5

OCR=2.5
OCR=4

R

3
2.5
2

a)

1.5
1

B=1

0.7
0.6
0.5
0.4
0.3

b)

0.2
0.1

0.5

0

0

0.2
0.1

0.2

0.3

0.4 0.5 0.6 0.7
Relative pressure

0.8

0.9

1

OCR=1.7~2.5

2.5

0.4

0.6

1

OCR=4

2

10
9

1.5

8
7

b)
R

1
0.5
0
0.2

0.4

0.6

0.8

B=2
B=1.5
B=1

6
5
4

a)

3
2

1

1
0

Relative Pressure

0.1

Fig. 2- a) Variation of the ratio of the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for four different soils,b) Variation of Ii with relative pressure

0.2

0.3

0.4

0.5

B=2
B=1.5

5

B=1

4
3
2

a)

1
0
0.2

0.3

0.4

0.5

0.6

0.7

Relative pressure

0.8

0.9

1

0.7

0.8

0.9

1

B=1.5

0.8

B=1

0.7
0.6
0.5
0.4
0.3
0.2

b)

0.1
0
0.2

0.4

0.6

0.8

1

Relative Preesure

Fig. 4- a) Variation of the ratio of the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for soil with OCR=1.7,b) Variation of β i with relative pressure
R

6

0.6

Relative pressure

0.9

Correction Factor

In order to define the correction factor of using
foundation width less than 2m in the immediate
settlement analysis, βi the ratio of the settlement
estimated from the Hardening Soil model over the one
estimated from elastic theory for foundation widths lessthan 2m is estimated and is divided by the results
obtained for the foundation assumed to have the width
equals to 2m. The results for normally and lightly
overconsolidated soils are shown in figs. 3 and 4.

R

0.8

Relative
Pressure
Fig. 3- a) Variation of
the ratioof the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for normallyconsolidated soil, b) Variation of βi with relative pressure

NC

3

Correction Factor

B=1.5

0.8

Correction Factor

4.5

R

The variation of βi doesn't differ for various soil types
and the graph introduced in fig. 4 can be used for soils
with other overconsolidation ratios.
Definition of the correction factors for consolidation
settlement is similar to those works done for immediate
settlement; for obtaining the correction factor of usingelastic theory in the consolidation settlement analysis,
Ic and the correction factor of using foundation width
less than 2m, βc the result of modeling with assuming
the soil to behave as the Hardening Soil model isdivided by those whom the elastic theory is used. The
results are smoothened and introduced in figs. 5, 6 and 7
for normally and lightly overconsolidated soils. Again
282

for other overconsolidation ratios, the extrapolation
should be employed.
b)

NC

5
4.5

OCR=1.7

4
3.5

OCR=2.5
OCR=4

R

3
2.5

Fig. 6- a) Variation of the ratio of the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for normallyconsolidated soil, b) Variation of β c with relative pressure

2

a)

1.5
1
0.5
0
0.1

0.2

0.3

0.4 0.5 0.6 0.7
Relative pressure

0.8

0.9

1
2.5

B=2

2.25

B=1.5
B=1

2

R

OCR=1.7~2.5

1.5
1.25

OCR=4

4

Correction Factor

1.75

NC

5

1

a)

0.75

3

0.5
0.1

2

0.2

0.3

b)

1
0

0.4 0.5 0.6 0.7
Relative pressure

0.8

0.9

1

1.2

0.2

0.4

0.6

0.8

B=1.5

1

B=1

1
Correction Factor

Relative Pressure

Fig. 5- a) Variation of the ratio of the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for four different soils,b) Variation of Ic with relative pressure

0.8
0.6
0.4

b)

0.2
0
0.2

0.4

0.6

0.8

1

Relative Pressure
B=2

9

B=1.5

8

Fig. 7- a) Variation of the ratio of the settlement estimatedfrom the Hardening Soil model over the one estimated fromelastic theory, R with relative pressure for soil with OCR=1.7,b) Variation of β i with relative pressure

B=1

6

a)

5

Similar to that mentioned about immediate settlement,for other soils with different overconsolidation ratios,
the variation of β c is not much different from the graphintroduced in fig. 7 and this graph also works for thesesoils as well.

4
3
0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1

Relative Pressure

5. Conclusion

2

Correction Factor

R

7

B=1.5
B=1

In this study, the methods introduced for estimating thesettlement in shallow foundations and their limitations
are considered in detail. It is concluded from priorstudies that the current introduced method should
consider the effects of both elastic and plastic strains

1.5

1

0.5

283

and secondary consolidation in analysis; the Hardening
Soil model which is used in PLAXIS software isconsidered and with the aid of this software and using
Hardening Soil model, a new relation is introduced for
estimation of settlement in shallow foundations which iscomposed of three components: Initial, consolidationand secondary consolidation settlements. The relation
also introduced the correction factors for Initial andsecondary consolidation settlements which are
estimated for four different clays with various
overconsolidation ratios.
References:[1] Buisman K, "Results of long duration settlement tests,"
Proceedings 1st International Conference on Soil
Mechanics and Foundation Engineering, Cambridge,
1936, Mass. Vol. 1, pp.103-107.
[2] Burland J.B, Broms B. B. and de Mello V. F. B.,"Behavior of foundations and structures", Proc. 9
ICSMFE, Tokyo, 1977, Vol. 2, pp. 495-546.
[3] Carter J. P., Booker J. R. and Small J. C., "The analysis
of finite elasto-plastic consolidation," Int. JNAMG,
1979, Vol. 3, pp. 107-129.
[4] D’Appolonia D. J., Poulos h. G. and Ladd C. C., "Initial
settlement of structures on clay," JSMFD, ASCE, 1971,
97(SM10), pp. 1359-1397.
[5] Davis E. H. and Polous H.G, "The use of elastic theory
for settlement prediction under three-dimentional
conditions," Geotechnique, 1968, 18(1), pp. 67-91.

284

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Numerical Study Of Earthquake Hazards Of Liquefaction Occurring On The Site
And Stability Of Transport Pipes
Hamed Bavanpouri, Dr.Jahangir Khazaie, Dr.Hasan Sharafi
Razi University, Kermanshah- Iran

ABSTRACT
Arteries are vital components of the buried pipe for water distribution, gas, oil, etc. are used. Vulnerability of pipelines against
earthquake-induced liquefaction has been observed since in many past earthquakes. There are extensive reports about.
Liquefaction can be by the side, fluid fragmentation, loss of cargo, buoyancy, causing subsidence and deformation in the
structure. Buried gas pipe network in some urban areas may be saturated loose sandy soil layers, are buried. The purpose of thisnumerical study on the liquefaction of the pipe is buried.
Keywords: liquefaction,buried pipelines,numerical analysis

1. INTRODUCTION
Better interaction between human and nature, requires
detailed understanding and knowledge of natural disasters.
Among natural disasters, more attention is needed toearthquake. Due to factors such as being unpredictable,
impossible to prevent the likelihood of being in crowded
places, in countries like Iran every year, several large andsmall earthquakes occur. It requires a comprehensive
protection plan is in various aspects. During an earthquake,
phenomena such as wave propagation and large movementsof the Earth, causing damage to the pipes. Great earth
movements such as fault movement, landslide and soilliquefaction of the most important factors is to count [1].
Studies on earthquake damage to buried pipes has provenabove ground stations on the structures, vibrations resulting
from seismic loads do not cause destruction of the buried
pipe. But the great movements of Earth's major pipe failure.
In other words, buried pipelines usually have the ability towithstand earthquake waves but The pipes are often unable towithstand the large motions of ground [2]. Large groundmotions can be caused by phenomena such as faults,
landslides and soil liquefaction. The main risk for pipelines
include earthquake ground motion and transient changes
lasting places of the earth. Transient ground motion caused
by waves emanating from the source, releasing energy andthe resonance is due to topography and local soil conditions.
Earth represents a lasting shift in the earth, defeat and failure
is caused by various phenomena such as : Surface faulting,the gable and landslide, lateral mass movement of soilliquefaction, the heterogeneous settlement. Research has
shown that most of the earthquake seismic damage to buried
steel pipelines in oil and gas by continuous change takes place
as a viable land and only a few of these injuries are caused by
the phenomenon of wave propagation. Psuedustatics steadyshift to enter the pipeline is not necessarily a lot of intensity

but however it may cause serious damage to the pipeline.
Example can be pointed to different earthquakes such as :Alaska and Nigita (1964), San Fernando (1971), Kobe
(1995), Chi Chi (1999) and etc. Soil liquefaction occurring in
Iran and also numerous harmful effects on arterial lines havebeen reported. Manjil earthquake that it can be mentioned[3].
Liquefaction effects on buried pipelines and proposedmodels (History Research)
Wang & Yeh [4] to the analysis of buried pipelines underliquefaction by Separating the two zones with liquefaction
and no liquefaction, a beam- column model is proposed forburied pipelines. In this modeling the geometric properties,
material specifications, soil stiffness, shear wave propgation
velocity, and axial bending effects, the mass and diameter, thebuoyancy effects of the soil mass was considered. The model
assumes the soil liquefaction around the environment, all lostits shear strength and behaves like a viscous fluid. Nishio[5]dynamic strain in the buried pipeline due to liquefaction ofsoil analyzes. The results show that under conditions of soilliquefaction, large strains are produced in the pipe. Also withregard to slip between soil and pipe strain on the pipeline tosignificantly reduced. Hamada and et al did Similar studies
with the large ground displacement caused byliquefaction-related chubu-Nihonkai earthquake in 1983.
The shear resistance between soil liquefaction severely
reduced re verberation in the sand [6].
Wang Guoxin and et al were examined seismic response ofpipelines buried in soil liquefaction caused by axial andlateralwaves
indifferent
directions[7].
In their model, the effective stress of soil and were floating ina spring system is used to simulate soil and pipe.

285

Fu-lu two methods to investigate the problem of staticliquefaction and its effects on buried pipelines will
provide[8].

2. Description of Numerical Model
2.1 Geometry of model
The data used in this modeling is based on information
related to gas pipelines in Kermanshah of Iran .
“TableІ. The characteristic parameters of the pipe”
LengthDiameter

Tichness

Poisson’s
ratio

Young’smodulus

12m

0.9m

0.1m

0.3

200e9 Pa

“Table П. The characteristic parameters of soil”“Fig. 2” Stress-strain curve of the pipe

Friction
angle

Dialation
angle

Density

Poisson’s
ratio

Young’smodulus

29º

2º

1700
kg/m^30.25

1e8 Pa2.3 Boundary conditions

In this study , Soil environment of the elastoplastic model isintended Drucker Pragr.

In this modeling, boundary conditions of the soil medium isconsidered infinite.The soil environment is assumed to besymmetric and isotropic. However, given the existing
conditions of the soil profile, it is fairly acceptable.

“Fig. 3” Boundary conditions of model“Fig. 1” Drucker Prager yield surface in 3D space
2.2 Specifications of Structual Pipe
As can be seen in Fig. 3 the stress - strain curve Ramberg
Osgood have used. the Ramberg Osgood relationship is :2.4 Pipe-soil ineraction
Interaction between soil and pipe for the consideration of thecontact elements are used .These elements, the contact forces
are two models :1- Vertical behavior is Hard contact. The first two levels, oneof which is usually more difficult level, is defined as areference and Another level it is defined as the function. Thisbehavior is assigned to the two surfaces in contact causes thepenetration of the surface elements of nodes in the elements286

of the reference level to prevent.
2- Tangential behavior that is related to the friction between
surfaces. Tangential to the friction coefficient of both fixedand variable coefficient model is available. The coefficient ofconstant friction, static friction to be considered , Thus
reducing the coefficient of friction between the sliding
surfaces are not considered But in the case of variable
coefficient of friction . reducing coefficient of friction due toslippage on the dynamic load can be made exponentiallydecreasing model .The coefficient of static to the dynamic
limit value is reduced. Represent the behavior of a weakened
area around the pipe [2] .
In this study, tangential behavior considered as constant.
In this case, the only parameter required is the coefficient offriction between surfaces .

Displacement(m)

-0.86
1

2

3

4

5

6

7

8

9

10

11 12

-0.88
-0.89
-0.9
-0.91
-0.92

Length(m)“Fig. 5” Ralationship between length of pipe anddisplacement
As can be seen in Fig. 6 With decreasing reduction ratio ofshear modulus, the excess pore water pressure reduced.

3 Numerical Results

Reduction shear modulus ratio

As can be seen in Fig. 4 Ralationship between excess porewater pressure ratio and strain of pipe due to liquefaction isshaped like a parabolic. With increasing excess pore waterpressure ratio, the strain of pipe is increased , but the strainreaches the maximum value and then decreases the amount ofstrain.

1
0.8
Strain-0.87 0

1
0.8
0.6
0.4
0.2
0
0

0.2

0.4

0.4

0.6

0.8

Excess pore water pressure

0.6

1“Fig. 6” Ralationship between excess pore waterpressure and reduction ratio of shear modulus

0.2
0
0

0.2

0.4

0.6

0.8

1

As can be seen in Fig. 7 With increasing excess pore waterpressure ratio, the buoyancy is also increasing

Excess pore water pressure ratio“Fig. 4” Ralationship between excess pore waterpressure ratio and strain

1

Buoyancy

0.8

As can be seen in Fig. 5 deformation due to liquefaction isshaped like a parabolic.

0.6
0.4
0.2
0
0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1

Excess pore water pressure“Fig. 7” Ralationship between excess pore waterpressure ratio and buoyancy287

4 CONCLUSION
Peak strains are caused when the excess pore water pressure
rises and falls.
The vibration strain peaks when the excess pore waterpressure ratio is between 0.45 to 0.55 during liquefaction
process. The longer the period of incomplete liquefaction
remains, the longer that of great vibration strain is.
The pipe strains shoud be classified according to themechanism of generation: vibration strains which indicate
pipe vibration, and accumulated residiual strains whichindicate the pipe bending due to uplift. Modes of pipe failure
indicated by the former is different from the later.
Given the significant changes that occur in the pipe, buried
pipe can be concluded that the gas network is vulnerable. And
retrofit them for prevention of human and financial losses, it
seems quite reasonable.
To obtain an equation to describe the pore water pressureduring the liquefaction phenomenon in this region requires
studies fare much worse.

5 REFERENCES[1] Liang,j.,Sun,S.” Site Effects on Seismic Behaviourof Pipeline”,Pressure Vessel Tech.,ASME,Vol .122,pp469-475,2000.
[2] Hoseini,M., Mehrantiv, “Seismic design guidelines foroil and gas pipelines” Tehran, International Institute of
Seismology and Earthquake Engineering,1997[3] “International Institute of Seismology and Earthquake
Engineeringguidelines for oil and gas pipelines” Manjil
earthquake ,Analytical Report Number 1 , 1990
[4] Wang L.R.L., Yeh, Y. H. ‘’Combined Effecte of Soil
Liquefaction and Ground Displacement to Buried
Pipelines’’. P. V. P., Vol.162,1989, PP.43-51.
[5] Nishio N. ‘’Dynamic Strains in Buried Pipelines due toSoil’’. P. V. P., Vol.162,1989,[6] Hamada M., Kubo K. and Saito K. ‘’Large Ground
Displacement and Buried Pipe Failure by Soil Liquefaction
during 1983 NIHONLAI-CHUBU Earthquake’’. P. V. P.,
Vol.98-4, 1985, PP.11-18.
[7] Guoxin Wang., Wei H. and Jianguo. ‘’Seismic Response
Liquefactional Conference on of Buried Pipelines During
Seismology and Earthquake Engineering, Tehran, Iran, Vol.2,
1995, PP.1343-1351.
[8] Fu-Lu M. ‘’Earthquake Response of Fluid-Filled
Pipeline in Soil’’. P. V. P., Vol.77, 1983, PP.196-201
288

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012

Analysis Of Geogrid Reinforced Soil Retaining Walls With Limit
Equilibrium And Finite Difference Method
Alireza sadeghabadi1 ,Jahangir khazaie2and Mostafa yousefi rad3
1

M.Sc. Student of civil Eng., Dept of civil Eng., Science & Research Branch.,Islamic Azad University,
Markazi(Arak),Iran; Tel: +98-9122070810; Fax:+982188532683;Email:αλιρεζα_σαδεγηι56≅ψαηοο.χοµ2Assistant Professor of Civil Eng., Department of Civil Eng., Univ. of Razi, Kermanshah, Iran;
Tel: +98-9123271926; Fax: +98-2144133580; Email: ϕ.κηαζαιε≅ραζι.αχ.ιρ3Department of earth Science., Payame Noor university.,Tehran,Iran; Tel: +98-9183638957;
Email:Ραδψουσεφι≅ψαηοο.χοµABSTRACTReinforced soil technologies have been widely applied during past few decades. A large variety ofreinforcing materials emerged and they have been developed for construction purposes, such as geogrids. In thisregards, due to increasing usage of polymeric materials such as geogrids shows the importance of investigating
on behavior of these types of materials. Geogrid reinforced retaining walls, are designed by experimental and theclassic methods which are often based on limit equilibrium methods. current analytical methods sometimes treat
the soil and reinforcement separately, which should been evaluated with practical situations. So, in this research
the experimental results compare with output results of finite difference method numerical analysis and limit
equilibrium methods in these types of walls. It's shown that the "F.D.M" results performed by FLAC2D softwareare in good agreement with those of experimental model tests and nearly manual calculations (L.E.M).

Keywords: geogrid, Geosynthetic, manual calculations, finite difference method “F.D.M", Limit equilibrium,
Polymeric reinforced soil retaining walls, FLAC2D software, laboratory models.

1.Introduction
Issued related to the design and factors affecting
the performance of reinforced soil have beenaddressed by many researchers in recent times(e.g. Bathurst et al.,2005;Park and Tan,2005;
skinnerandRowe,2005;Versuo
etal.,2005;Hufenus et al., 2006;Nouri et al.,2006).
Also, the behavior of reinforced earth structures
has been comprehensively studied through filed
observation of full scale physical model,laboratory model testing, and numericalsolutions. However the cost of constructing andmonitoring
full
scale
reinforced
test
embankments is quiet high. An alternative
method such as a numerical “experiment” orsimulation by means of appropriate methods
such as finite difference (F.D.M) or limit
equilibrium(L.E)
techniques(e.g.and
Rowe,1994) is essentially required.
So, numerical modeling of reinforced systems
behavior is important for the evaluation of

Knowledge concerning these systems, theirdesign methodologies and also to forecast their
behavior. with this goal, a real soil wall withlateral slopes reinforced with geogrids wasnumerically modeled. The main target of thisresearch is to analyze the influence ofreinforcement and the reinforcement geometry inthe behavior of soil systems reinforced withgeogrids.

of extension(Mendonca 2004).

The maximum inclination of the reinforced slope

2.Reinforced soil Description
A soil wall with slopes reinforced with geogrids
was built between Regua and Reconcos in themain itinerary 3 (IP3) and is a part of thereestablishment 2. The reinforced slope has anextension of about 206m, is in curve and thereinforced area reaches a maximum height ofabout 20 m in the outside curve slope at 150 m

289

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012is about 60o in horizontal direction. On thereinforced sections higher than 10m, withvariable width parallel to the road pavement was
built. The maximum width of the bench was 3 m.
the bench had the following objectives: to reduceminimum inclination of the slope in the zones ofthe larger height, to facilitate drainage and,
consequently, to reduce the superficial erosion.

in the fill and are presented in Table (1). This soil isa residual granite and the layers were placed with aheight of 0.30 m after compaction.
3. Numerical modeling of the reinforced soil wall.
for this work the program FLAC V4.0 was chosen.
The main reason for this decision was the fact that it
is commercial software and so it will be easier for
geotechnical engineers’ to access the code. The
construction the soil wall was modeled considering
the real geometry, the adopted sequence construction,
the materials used and the real solicitations applied.
The total number of was 1700, with 1640 square twodimensional elements representative of the soilmaterial, 30 elements of the type representative of theface and 30 linear cable elements representative of
the geogrids(figure.1)

The simulation of the soil construction was modeled
as a successive placement of 30 soil
layers and equal number of geogrids levels. So, thetotal number of calculation phases was 30. The soil
was doubted with elastic-plastic behavior with afailure criterion of Mohrcoloumb. Geosynthetic
layers were modeled using linear, elatic-plastic cableelements.
The face characteristics were modeled consideringin a strip of 0.5m adjacent to the face soil with acohesion equal to 25 KN/m2. The reinforcements
are placed horizontallyand spaced 0.60 m
vertically are high-density polyethylene (HDPE)
uniaxial geogrids with tensile strengths selectedaccording to the height of the soil wall.
The face units, 0.60 m high, were constructed using
welded wire net with quadrangular 0.15m size
openings. The face unit angle (with the horizontal) is
60º.
The characteristics of the fill material used in theslope nearby the facing were similar than that used

Figure(1). Mesh elements of the embankment
numerical model.

Table (1). Specifications of the soil.

290

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012

From the results of the numericalmodeling presented in Figures 2 & 3 it wasverified that, only in a reduced extension to
the interior of the embankment face,displacements
haveinfluenceon
geogrids’s deformations and in consequenceon its tensile load distribution. Therefore,
we think that it is interesting to study theeffect of position and reinforcement length
in the limitation of the slope deformations.

Even so with two different lengths: main
reinforcements spaced 1.2 m, with the length
considered in the base model; and secondary
reinforcements, spaced 1.2 m in the middleof the main reinforcements, with variable
lengths of 8,6, 4 and 2 m. Additionally, acalculation was performed with thereinforcements spaced vertically at 1.2 m.

So, starting from a calculation withreinforcements with the same length (12.5m) designated by base model, several
calculations were made. In these calculations
the number of reinforcements was the sameas the base model.

Figure(3). Contours of horizontal displacements.

The Conclusions from N.M. show that the
maximum horizontal face displacements arestrongly dependent on the length of the intermediate
reinforcements. Figure 4 clearly shows that the facedisplacements decrease with the reduction of thelength of the intermediate reinforcements until valuesof about 4m. For shorter reinforcements, thedisplacements in-crease significantly. Therefore, theoptimum intermediate reinforcement’s length, forthis particular structure, is between about 4 and 6 m.

Figure (2). Geogrids displacements

This situation results from the fact that thereinforced soil wall, with shorter intermediate
reinforcements for the optimum values, are flexible
in a non-uniform way and consequently the
resulting rotation due to differences of stiffness
along the reinforced Soil drive to a reduction of
the horizontal deformation near the face. In thebase model, as in models
with
short
intermediate reinforcement length, the stiffness ofthe soil mass is more uniform and consequently the291

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012

horizontal face deformation, in the absence of anaccentuated rotation, is controlled by the stabilityof the face. In limit, as it would be expected, its
maximum deformation occurs when intermediate

reinforcements do not exist.

Figure(4). Effect of secondary Georid’s length on
maximum horizontal face displacements.
middleverticaldistancefromprimary
reinforcements, spaced in the vertical of 1.2m(Fig.5)

Applying this method to the IP3 embankment
with reinforced slopes with geogrids, an alternative
solution to the real structure could be thecorresponding optimized calculation. That is,secondary reinforcements (with 4.0 m of length), at

Figure 5. Normalized geogrids distribution for
the geogrid reinforced soil retaining wall.

Figures (6) shows some numerical results and can becompared with those presented in Figure (3). From
its analysis it is verified that the influence of the facedisplacements on the interior of the reinforced zone
increases in the Normalized solution.

292

GEOMAT 2012
Second int'l Conference on Geotechnique,
Construction materials and Environment.
14-16 November 2012

Figure (6). Horizontal displacements of the Normalized model.

4.Conclusions
Although, Internal stability analysis geogrid
reinforced soil retaining walls, based on limit
equilibrium, are known to be very conservative
current design methods do not provide usefulperformance information such as wall face
deformations.

reviewing the main results of the N.M. in static
conditions of the reinforced soil with geogrids built
in IP3, between Regua and Reconcos, it’s shownthat the control of face deformations is guaranteed
considering geogrids with an minimum length plus a
sufficient number of long geogrids to ensure thestructure’s internal and external stability.
It’s observed that in a normalized geogrids solution
the amount of shorter geogrid’s length should have
between 30-50% of the long geogrid’s length.

Utilizing the constitutive models and properties offoundation soils published by previous researches,
numerical simulations were conducted using 2D
explicit Finite Difference programs, FLAC2D. after
4. Refrences[1]Mendonça, A.A. 2004. Study of the behaviour ofgeogrids re-inforced structures. Ph.D. thesis, FEUP,
Porto, Portugal, (in Portuguese)[2] Rowe, R.K. and skinner, G.D .(2001) Numericalanalysis of geosynthetic reinforced retaining wallconstructed on a layered soil foundation, Geotextile
and Geo-membranes, 19(7), 387-412.
[3] ITASCA Consulting Group, Inc. 2000. FLAC –Fast Lagrangian Analysis of Continua, Ver. 4.0
User’s Manual. Minneapolis: Itasca.

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Recent Development In Electrokinetic Stabilization Of Soft Soil: A Review
Hossein Moayedi1, Sina Kazemian2, and Bujang B. K. Huat3,
1. Department of Civil Engineering, Science and Research Branch, Islamic Azad University, Tehran, Iran.
2. Department of Civil Engineering, Bojnourd Branch, Islamic Azad University, Bojnourd, Iran.
3. Department of Civil Engineering, Universiti Putra Malaysia, Serdang, Selangor, Malaysia.
ABSTRACTElectrokinetic (EK) stabilization is a ground improvement method which is treats soils without excavation process. Thisis known as an important advantage over traditional grouting methods. It has been used in several soil improvement
categories including consolidation, contaminant removal, dewatering, and stabilization. In addition, the EK
applications cause electrochemical effects in the treated soil, leading to changes in the soil’s chemical, physical, andmechanical properties. With EK injection technique, chemical stabilizing agents are electrically injected into the soils toachieve stabilization by means of ion exchange, precipitation or mineralization. The aim of this study was to review the
recent development in the EK strengthening of soft soils. Enhancing the EK stabilization techniques require a deeply
assessment through the colloidal surface charge and advances in the EK apparatus.
Keywords: Electrokinetic, Soil improvement, Grouting, Direct current, Soft soil.

1. INTRODUCTION
Soil improvement refers to any method or techniques that
improve the engineering properties of soil, likecompressibility, shear strength, stiffness, and permeability.
Raju (2009) classified the soil improvement methods tofollowing principles: (i) chemical modification (e.g. jet
grouting, deep soil mixing, injection grouting), (ii)consolidation (e.g. vacuum consolidation, prefabricatedvertical drains & surcharge, stone columns), (iii)
reinforcement (e.g. geosynthetic reinforcement, stone
columns), and (iv) densification (e.g. dynamic compaction,
vibro-compaction, compaction grouting) [1, 2] (Table 1).
Majority of the companies and geo-engineers, which are
working in these fields agree that the soil improvement
methods are typically done based on followingclassifications: (i) hydraulic modifications, (ii) mechanicalmodifications, (iii) modifications by inclusions andconfinement, and (iv) physical and chemical modifications[3]. The goals of such improvement methods include
reducing soil water content, reducing compressibility, andincreasing shear strength.
2. DEFINITION OF ELECTROKINETIC
EK is well known as the physicochemical transport of charge,
action of charged particles, and effects of applied electric
potentials on formation and fluid transport in porous media[4, 5]. As stated, the presence of the diffuse double layer(DDL) causes several EK phenomena in soil, which may
result from either the movement of different phases in respectof each other including transport of charge, or the movementof different phases relative to each other due to the applyingof such electric potentials [4, 6]. Besides, application of direct
current (DC) through electrodes give rises to electrolysis
reactions at the electrodes [7, 8]. Oxidation of water at theanode generates an acid front and reduction at the cathodegenerates a base front. Electrolysis reactions are described by
the Equations (1) and (2).
2H2O – 4e- → O2 ↑ + 4H+ → (anode)
4H2O – 4e- → 2H2 ↑ + 4OH- → (cathode)

..(1)
(2)

Within the first few days of EK processing, electrolysis
reactions increase the pH at the vicinity of the cathode anddecrease the pH at the vicinity of the anode. These changes
are normally dependent to the total current applied [5, 9].
The EK phenomena include electroosmosis, electrophoresis,
streaming
potential,
andsedimentation
potential.
Electroosmosis is defined as movement of fluid with respect
to a solid wall as a result of an applied electric potential
gradient [10, 11].
In other words, if the soil is placed between two electrodes ina fluid and an electromotive force is applied the fluid will
move from one side to the other. Electrophoresis is the
movement of suspended solid in a liquid due to application ofan electric potential gradient. Streaming potential is thereverse of electroosmosis. It defines the generation of anelectric potential difference due to fluid flow in soils.
Sedimentation potential (or migration potential), known as
Dorn effect [12], is an electric potential generated by the
movement of particles suspended in a liquid [9, 13, 14].

294

Table 1. Various soil improvement techniques (after Hausman, 1990)
Smooth Steel Rollers andPneumatic–tired Roller
Static Roller
Sheepsfoot Rollers
Shallow Surface
Compaction
Impact and Vibratory
Equipment

Grid Roller
Tamper, Rammers, and Plate
Compactors
Vibrating Rollers
Impact Rollers

Mechanicalmodifications

Pre–compression
Deep Compaction
Technique

Explosion
Heavy Tamping (Dynamic compaction)
Vibration (Vibro–compaction and Vibro–replacement)

Hydro–mechanical
Compaction

Hydraulic Fill
Dry Fill with Subsequent Spraying or Flooding
Compaction of Rock Fill with Water Jets

Open Sumps and Ditches
Hydraulicmodify cations

Gravity Flow Wells
Pre-loading and Use of Vertical Drains
Vacuum Dewatering Wells and Preloading
Electrokinetic Dewatering and Stabilization
Mixing Method

Physical andchemicalmodifications

Shallow Mixing Method (SMM)
Deep Mixing Method (DMM)
Permeation Grouting

Injection and grouting

Compaction Grouting
Hydro fracture Grouting
Compensation Grouting

Flexible Geosynthetic Sheet Reinforcement
Modifications by
inclusions
and confinement

In Situ Ground
Reinforcement
Soil Confinement by
Formwork

Ground Anchorage
Rock Bolting
Soil Nailing
Crib Walls
Gabions and Mattress
Fabric Formwork

3. ELECTROKINETIC HARDENING OF SOIL
While conventional soil grouting can be used to stabilize soils
with relatively high permeability (k typically greater than 10-5m/s), other methods need to be applied for strength
improvement of fine-grained soils with lower hydraulicconductivity values. These include, hydrofracture grouting,artificial ground freezing, and induced consolidation either byapplying electric fields or surcharging the soil [15]. Although
all of these methods result in some considerable degree ofground movement, it still have other drawbacks [8, 15]. With
the increase in urban construction and construction inmechanically substandard soils, new methods was required tostabilize soft soils with low permeability while minimizing
ground movements [11]. EK stabilization is a ground
improvement method which is treats soils without excavation(e.g. disturbing the soil). This is known as an important
advantage over traditional methods. It has been used inseveral soil improvement categories including consolidation,
contaminant removal, dewatering, and stabilization [7, 16-20]

295

3.1 Electroosmosis Consolidation
Early application of electroosmosis was intended for
dewatering of wet soils and soil consolidation [21-23].
Enhanced electroosmosis with injection of cationic stabilizer
into the soil has been developed significantly in recent years
to improve the shear strength of soft soils [14, 15, 24-27]. The
enhanced electroosmosis techniques are different thanelectroosmotic consolidation in that transport processes are
utilized for injecting chemical grouts into the soil.

EMs separate positively and negatively charged ions andcause their migration to opposite electrodes. As a result,hydraulic flow might improve the migration of certain ions,but retard migration of other ions (with opposite charge) [5, 9,31]. The relative contribution of electroosmosis andmigration to ion transport under electric fields varies fordifferent soil types, type of ion, water content, boundary
conditions, and pore fluid chemistry.

Gray and Schlocker (1969) and Gray (1970) studied
electroosmosis as a transport mechanism for hardening soft
soils by injecting the aluminum ions [28, 29]. In their studies,
the liquid limit of montmorillonite dropped to about half of
the initial value. Sutton and Alexander (1987) described using
electroosmosis to draw chemicals from anode to cathode forsoil stabilization, further indicating that it is possible to alter
the clay properties and cause mineralization byelectrochemical methods [8]. Shang et al. (2004) statedconditions that most favored for a chemical reagents used in
the electrochemical cementation process are (i) the chemicalcompounds should increase the soil shear strength, (ii) theymust be economically feasible, and (iii) they should besoluble in water so that they can be transported by theelectrical current through soil pores.

4. IMPORTANCE OF SURFACE CHARGE

While electroosmosis mobilizes the pore fluid, carrying
soluble grouts toward the cathode, ionic migration effectivelytransports negative anions to the positive anode and positive
cations to the negative cathode. The important difference isthat this mode of chemical transport can potentially occur
without any fluid flow, compared to the use of electroosmosis
to induce chemical transport [22]. Based on this fact, Ozkan
et al. (1999) presented the results of experiments for injectionof phosphate and aluminum ions at the cathode and anode to
improve the mechanical properties of kaolinite [30]. The
premise for this work is that electroosmosis can carryphosphate ions towards the cathode while ionic migrationtransports phosphate ions towards the anode, resulting in
more rapid, uniform soil stabilization. The results showed a
500–600% increase in kaolinite undrained shear strength after
treatment. However, the strength increase was nothomogeneous throughout the specimen. Previous laboratory
work and case studies are summarized in Table 2.

Surface charge in turn is a function of pH and dissolved salts
concentration which normally measure by the zeta (ζ)
potential [38, 39]. The ζ potential is known as an intrinsic
property of colloidal particles suspended in a liquid.
Adsorption of ions or dipolar molecules is determined by, andalso determines, this charge and potential distribution [40,41]. The potential distribution itself determines theinteraction energy between the particles, and this is in manycases responsible for the stability of particles towards
coagulation and for many aspects of the flow behavior of thecolloidal suspension [27, 42, 43].
5. CONCLUSION
Although EK injection technology has been seldom
employed, as showed in Table 2, the majorities ofapplications have been successful and have proved an
economic alternative to traditional ground improvement
methods [14, 24, 30]. Over the last 40 years, a few hundred
different compounds of chemical grout are introduced. The
grouting and chemical grouting technologies have also grownover the last few decades. Besides, the EK injection isestablished as an effective in situ ground improvement
technique in fine soils, particularly where ground movements
need to be minimized. However, the origin of EK treatment
performed on a few soil types as well as chemical stabilizers.
For most of the soft soil, EK injection phenomena andconsequences of injection of cationic (or chemical stabilizer)
reagents have not been studied yet.

3.2 Application of Electrokinetics in Soil Improvement
Alshawabkeh (2009) stated that application of electric
gradients in soil mass will result in two important transportmechanisms; electroosmosis as well as electro ion migration(EM). Electroosmosis draws chemical stabilizers with theflowing water under electric fields. However, the EMtransports ions to the electrode opposite in polarity underelectric fields [9]. Electroosmosis and any other hydraulicflow will usually carry and move all types of solutes from one
location to another, depending on flow direction. However,296

Table 2. Various reported case studies and researches relating to EK stabilization
Reference
Application
Soil
Stabilizer
Comments[32]
Foundation forSoft to stiff
Sodium silicate
Major implementation
bridgeclay
problems
abutment
encountered[33]
Laboratory
Liquifiable
Silicate solutions,
Successful application
investigation
sand
bentonite,
aluminum hydroxide[34]
Highway
Low-plasticity Potassium chloride
Reductions in degree of swell
subgrade,clay
and swell
Arizonapressure[35]
Foundation
Loess
Sodium silicate
Successful application
strengthening[5]
Large scale
Fine grained
ammonium at
Successful applicationlaboratory
sand bed andcathode and sulphateinvestigation
Kaolinite bedat the anode[36]
Laboratory
Kaolinite
Aluminum andLarger increases in shear
investigation
phosphate ionsstrength for phosphoric acidthan aluminum
sulphate/phosphoric acid
[30]
Laboratory
Kaolinite
simultaneously
average shear strength increase
investigation
inject aluminum ions of 500 – 600% was obtained inat the anode andthe region of the cathodephosphoric ions at
the cathode[37]
Laboratory
Sand
Sodium silicate andPossibility that temperatureinvestigation
calciumvariation in ground underchlorideelectric loading has an influenceon strength of improved ground.
[15]
Laboratory
Marine illitic
inject phosphoric
shear strength increased byinvestigationsoil (Boston
acid at the cathode160% at the side of cathodeBlue Clay)
6. REFERENCES[1] Raju, VR, Ground Improvement Principles andApplications in Asia. Ground Improvement Technologies
and Case Histories, 2009, pp. 43-66.
[2] Kazemian, S, Stabilization of Peat by Cement-Sodium
Silicate Grout Using Injection-Vacuum Technology.
Department of Civil Engineering, Doctor of Philosophy,
2011, pp. 267.
[3] Hausmann, MR, Engineering Principles of Ground
Modifications. 1990.
[4] Acar, YB, AN Alshawabkeh, RJ Gale, "Fundamentals ofExtracting Species From Soils by Electrokinetics," Waste
Management, 13(2), 1993, pp. 141-151.
[5] Acar., YB, MF Rabbi., EE Ozsu., "Electrokinetic
Injection of Ammonium and Sulfate Ions Into Sand andKaolinite Beds," Journal of Geotechnical andGeoenvironmental Engineering,
123(3), 1997, pp.
239-249.
[6] Acar,
YB,
AN
Alshawabkeh,"Electrokinetic
Remediation. I: Pilot-Scale Tests With Lead-Spiked
Kaolinite,"
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GeotechnicalandGeoenvironmental Engineering,
122(3), 1996, pp.

173-185.
[7] Asavadorndeja,
P,
U
Glawe,"Electrokinetic
Strengthening of Soft Clay Using the Anode
Depolarization Method," Bulletin of Engineering
Geology and the Environment, 64(3), 2005, pp. 237-245.
[8] Alshawabkeh, AN, TC Sheahan, X Wu, "Coupling ofElectrochemical and Mechanical Processes in Soils Under
DC Fields," Mechanics of Materials, 36(5-6), 2004, pp.
453-465.
[9] Alshawabkeh, AN, "Electrokinetic Soil Remediation:
Challenges and Opportunities," Separation Science andTechnology, 44(10), 2009, pp. 2171-2187.
[10] Mitchell, JK, K Soga, Fundamentals of Soil Behavior.
2005.
[11] Acar, YB, RJ Gale, AN Alshawabkeh, RE Marks, S
Puppala, M Bricka, R Parker, "Electrokinetic
Remediation: Basics and Technology Status," Journal ofHazardous Materials, 40(2), 1995, pp. 117-137.
[12] Kruyt, HR, Colloid Science (I): Irreversible Systems,.
1952.
[13] Acar, YB, RJ Gale, GA Putnam, J Hamed, RL Wong,"Electrochemical Processing of Soils: Theory of pH
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[14] Chien, SC, CY Ou, MK Wang, "Injection of Saline
Solutions to Improve the Electro-Osmotic Pressure andConsolidation of Foundation Soil," Applied Clay Science,
44(3-4), 2009, pp. 218-224.
[15] Alshawabkeh, AN, TC Sheahan, "Soft Soil Stabilization
by Ionic Injection Under Electric Fields," Ground
Improvement, 7(4), 2003, pp. 177-185.
[16] Casagrande,
L,"Electro-osmosis
inSoils,"
Geotechnique, 1(3), 1949, pp. 159-177.
[17] Azzam, R, W Oey, "The Utilization of Electrokinetics inGeotechnical and Environmental Engineering," Transport
in Porous Media, 42(3), 2001, pp. 293-314.
[18] Han, SJ, SS Kim, BI Kim, "Electroosmosis and Pore
Pressure Development Characteristics in Lead
Contaminated Soil during Electrokinetic Remediation,"
Geosciences Journal, 8(1), 2004, pp. 85-93.
[19] Kim, SO, WS Kim, KW Kim, "Evaluation ofElectrokinetic Remediation of Arsenic-ContaminatedSoils," Environmental Geochemistry and Health, 27(5-6),
2005, pp. 443-453.
[20] Nieto Castillo, AM, JJ Soriano, RA García-Delgado,"Changes in Chromium Distribution During theElectrodialytic Remediation of a Cr (VI)-ContaminatedSoil," Environmental Geochemistry and Health, 30(2),2008, pp. 153-157.
[21] Bjerrum, L, J Moum, O Eide, "Application ofElectro-osmosis to a Foundation Problem in a Norwegian
Quick Clay," Geotechnique, 17(3), 1967, pp. 214-235.
[22] Gray, DH, JK Mitchell, "Fundamental Aspects ofElectroosmosis in Soils," Journal of the Soil Mechanics
and Foundations Division, ASCE, 93(SM6), 1967, pp.
209-236.
[23] Esrig, MI, "Pore Pressures, Consolidation andElectrokinetics," J. Soil Mech. Found. Div., Am. Soc. Civ.
Eng, 94(4 SM), 1968, pp. 899-921.
[24] Lefebvre, G, F Burnotte, "Improvements ofElectroosmotic Consolidation of Soft Clays by
Minimizing Power Loss at Electrodes," Canadian
Geotechnical Journal, 39(2), 2002, pp. 399-408.
[25] Abdullah, WS, AM Al-Abadi, "Cationic-Electrokinetic
Improvement of an Expansive Soil," Applied Clay
Science, 47(3-4), 2010, pp. 343-350.
[26] Ou, CY, SC Chien, HH Chang, "Soil Improvement
Using Electroosmosis With the Injection of Chemical
Solutions: Field Tests," Canadian Geotechnical Journal,
46(6), 2009, pp. 727-733.
[27] Ou, CY, SC Chien, YG Wang, "On The Enhancement ofElectroosmotic Soil Improvement by the Injection ofSaline Solutions," Applied Clay Science, 44(1-2), 2009,
pp. 130-136.
[28] Gray, DH, "Electrochemical Hardening of Clay Soils,"
Geotechnique, 20(1), 1970, pp. 81-93.
[29] Gray, DH, J Schlocker, "Electrochemical Alteration of

Clay Soils," Clays and Clay Minerals, 17, 1969, pp.
309-322.
[30] Ozkan, S, RJ Gale, RK Seals, "Electrokinetic
Stabilization of Kaolinite by Injection of Al3+ and PO 43 Ions," Ground Improvement, 3(4), 1999, pp. 135-144.
[31] Park, JY, Y Chen, J Chen, JW Yang, "Removal of
Phenanthrene From Soil by Addidive-Enhanced
Electrokinetics," Geoscience Journal, 6(1), 2002, pp. 1-5.
[32] Caron, C, "Consolidation of Clayey Soils by Using
Electro-Osmosis, (Consolidation Des Terrains Argileux
par Electro-Osmose)," Ann Inst Tech Batiment Trav Pub,(285), 1971, pp. 73-123.
[33] Yamanouchi, T, S Matsuda, 'Proceedings of the
Symposium on Recent Developments in the Analysis ofSoil Behaviour,' 1975, pp. 411-421.
[34] O'Bannon,
CE,
GR
Morris,FP
Mancini,"Electrochemical Hardening of Expansive Clays,"
Transportation Research Record,, (593), 1976, pp. 46-50.
[35] Oncescu, V, RJ Bailie, "Problems of Soil Consolidationand Waterproofing Solved by Grouting," Studii de
Cerictare in Constructi Hidraulice Si Mechanica Rocii,
19, 1977, pp. 127-210.
[36] Anon, "Chemical Stabilisation of Kaolinite byElectrochemical Injection," Geotechnical Special
Publication, 81(286-297), 1998.
[37] Fujihira, M, E Shimizu, T Konishi, S Yasuda, T Shimizu,"The Fundamental Study on Soil Prevention Method ofLiquefaction Using Electrophoresis of Fixative
Materials," Zairyo/Journal of the Society of Materials
Science, Japan, 49(1), 2000, pp. 38-41.
[38] Brunelle, MT, "Colloidal Fouling of Reverse Osmosis
Membranes,," Desalination, 32, 1980, pp. 127-135.
[39] Niriella, D, RP Carnahan, "Comparison Study of Zeta
Potential Values of Bentonite in Salt Solutions," Journalof Dispersion Science and Technology, 27(1), 2006, pp.
123-131.
[40] Hlavacek, M, JF Remy, "Simple Relationships among
Zeta Potential, Particle Size Distribution, and Cake
Specific Resistance for Colloid Suspensions Coagulatedwith Ferric Chloride," Separation Science andTechnology, 30(4), 1995, pp. 549-563.
[41] Sabah, E, U Mart, M Çınar, MS Çelik, "Zeta Potentials
of Sepiolite Suspensions in Concentrated MonovalentElectrolytes," Separation Science and Technology,
42(10), 2007, pp. 2275-2288.
[42] Carlson, JJ, SK Kawatra, "Factors Affecting Zeta
Potential of Iron Oxides," Mineral Processing andExtractive Metallurgy Review, 2011.
[43] Gallardo., MAM, MR Calzado., MML González., GC
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298

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Application of Ion-Exchange Solution Soil Stabilizer in Increasing
CBR Values of Clayey Soils
Reza Ziaie Moayed1, Seyed Ali Hashemi2 and Farzad Allahyari3
1,3Imam Khomeini International University, Iran
2
Iran Polymer and Petrochemical Institute, Iran
P

P

P

P

P

P

P

ABSTRACT
Soil stabilization is one of the methods for increasing soil strength. For this purpose many of materials can be used like
lime, fly ash, cement and organic polymers. Using this material has some limitation and does not improve soil properties
very well. One of the best materials for soil stabilization is ion-exchange solutions. There are many ion-exchangesolutions soil stabilizer and one of the best and the most economical ion-exchange solution is using in this paper and its
name is CBR+4 . This paper describes the application of ion-exchange solution CBR+4 (CBR plus) for stabilization two
various clayey soils. California Bearing Ratio (CBR) tests were carried out on unstabilized and stabilized soil samples.
The samples were soaked in water for 96 hours. CBR values of stabilized soils in the soils with lower plasticity index are
more than unstabilized. For soils with lower plasticity index CBR values increases up to the 168%.
P

P

P

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Keywords: Soil stabilization, ion-exchange solution, CBR+4, CBR, Atterberg limitsPP

1. INTRODUCTION
Clay soils make up great part of soils in the ground. Clayeysoils are commonly stiff in dry states but lose their stiffness
when they are saturated with water. Clayey soils commonly
have low bearing capacity, low stability, high settlement, andexcessive swelling or shrinkage properties. For these reasons
it is necessary to make such soils suitable for construction.
The use of traditional geotechnical engineering techniques forinfrastructure, such as the replacement of unsuitable soils forstiff and resistant embankment, is often problematic, not only
for their high costs, but even more for environmental reasons.
In roads, for instance, the use of granular bases becomesunsuitable when the extraction site is at a significant distancefrom the construction site [1]. Application of stabilization
agent on soils has a long history. Many materials such as lime[2], fly ash [3] and organic polymers [4] were used asstabilizing agents. When lime is added to a clayey soil it has
an immediate effect on the properties of the soil. With adding
lime cation exchange begins to take place between themetallic ions associated with the surfaces of the clay particles
and the calcium ions of the lime. One of the ways to makeclayey soils suitable for construction is altering the properties
of soil. This increases the strength, reducing compressibility,
swelling or shrinkage and increasing the durability of soils.
Ion exchange is one of the methods for stabilization of clayeysoils. It is well-known that the swelling properties of
expansive soils significantly are affected by cation exchange
capacity [5]. Ion exchange occurs when some of the addictive
such as fly ash, lime, cement and ion exchange solution areadded to these soils. Many of ranchers have investigated theimprovement of the expansive soils with fly ash and they

have determined that addition of fly ash to the soil decreases
the value of cation exchange capacity [6], [7].
There are many ion-exchange solutions soil stabilizer and oneof the best and most economical ion-exchange solution isusing in this paper and its name is CBR+4.
The objective of this paper is to investigate the bearing ratio
capacity of two clayey soils that are stabilized with CBR+4 ion
exchange solution. The tests were carried out on two clayeysoils with different doses of CBR+4 (0.0, 0.12 and 0.28
(gr/lit)) in two different curing times. The test results also
were discussed in detail and compared with each other.
P

P

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2. MATERIALS
2.1 SoilIn this experimental study two fine-grained clays were used.
Soils are from Iran- Qazvin- Abyek. The grain size
distributions of two clayey soils are presented in Fig.1.
The soils lie above the A-line in plasticity chart. Two soils areclassified as low plasticity soil according to the unified soil
classification system (ASTM D422-87) and their nameaccording to USCS is CL (clayey soil with low plasticity).
The optimum moisture contents and maximum dry densitiesof two clayey soils were obtained according to ASTM D4318(Fig. 2) .Engineering properties of two soils are presented inTable 1.
Soil samples for CBR test were compacted with theiroptimum moisture contents value and the respective dry
densities.

299

the soil [9]. Table 2 shows physical and chemical propertiesof CBR+4.
P

P

Table 1: Engineering properties of two clayey soils
property
Atterberg
limitsCompaction
parameters

soil Asoil B

Liquid Limits, (%)

30

42

Plastic Limits, (%)

18

23

Plasticity Index, (%)
Optimum Moisture
Content, (%)

12

19

15

17

19

17.8

CL

CL

Maximum dry unit
weight,(kN/m3)
P

Fig.1 Grain size distribution of two clayey soils

P

Soil classification(USCS)

Table 2: Physical and chemical properties of CBR+4 [10]Chocolate Brown viscousAppearance
fluid
Odour
Sulphurous OdourP

Fig.2 Compaction curves of two clayey soils
2.2 CBR+4 Solution
CBR+4 solution is a stabilizer for anionic soils that can
improve the bearing capacity of a variety of clayey soils. Soils
that normally are not suitable for road construction can now
be treated with CBR+4 for base and subbase in road
construction. Clay minerals inherently have a predominantly
negative electrical or anionic charge; this causes the clay
minerals to have strong attraction for any caution that arepresented. These negative clay minerals attract cautions likeiron filings to a magnet and will react with water whenpresent. Normal temperatures and compaction pressure will
not remove them. This layer of water is known as theelectrostatic diffused double layer of adsorbed water [8].
Generally it can be said that CBR+4 forms an extremely thinoily layer on the surface of soil particles and especially onclay particles. This facilitates the compaction of soil andallows water, which is normally chemically bound with thesoil particles, to be driven out of the soil matrix. In thisprocess the soil can be compacted to a higher density. Since
CBR+4 neutralizes the natural electrical charges that occur on
the soil particle surface, the soil particles can be compacted toa much closer degree (especially by traffic forces). This willlead to increasing in internal friction between the soils
particles, which in turn result in a higher bearing capacity for
P

P

P

P

Physical State

Viscous Fluid

Freezing Point(°C)<-10° C

Boiling Pint (°C)

100°C

pH

9

Specific Gravity

0.94

Coefficient Water/Oil

100% water soluble

Percent Soluble (at 20°C)

100%

P

3 EXPERIMENTAL PROGRAM
3.1 Test Procedure
To study the effect of various amount of ion exchange CBR+4
and curing time on the bearing ratio of clayey soils withvarious plasticity index(PI) , a series of bearing ratio tests
were conducted on soaked samples for three different amountof CBR+4 solution (without CBR+4 and two different doses ofit) and two different curing times. For soaked condition, thesamples were immersed in water for 96 hours. The purpose ofdifferent amount of ion exchange solution and curing time is
to investigate the effect of each variable on bearing ratio ofdifferent soil samples.
P

P

P

P

P

P

3.2 Preparation of Samples
The bearing ratio mold is a rigid metallic cylinder with an
inner diameter of 152mm and a height of 178mm that 61mm
of its height is filled with a circular metal disk.
To prepare the samples, oven–dried clayey soils were mixed
with their optimum moisture content (soil A with 15%
moisture and soil B with 17% moisture).The soil that was
mixed with its optimum moisture was placed in five layers at
the bottom of the mold. Each layer was compacted by 56
blows of a 44.5-N rammer dropped from a height of 457mm.

300

The samples with various clayey soils and amount of CBR+4
solution have been cured for two various times, one day and
one week. After these curing times, samples were soaked inwater for 96 hours.
To prepare the soaked samples, the mold containing clayeysoil was immersed in water, allowing free access of water tothe top and bottom of the samples, and was allowed to soakfor 96 hours (ASTM D1883-99).

deteriorate its strength. But soil B shows reduction in CBRvalues with increasing the dose of CBR+4.

3.3 California Bearing Ratio (CBR) Test
The bearing ratio is one of the fundamental parameters used
to demonstrate the behaviour of soil. This test is used in many
geotechnical projects including roads, railroads, pavements
and airport runways. To demonstrate the influence ofdifferent variables on the bearing ratio of the clayey soil
stabilized with CBR+4 solution, a series of bearing ratio tests
were carried out on stabilized and unstabilized specimens.
The bearing ratio tests were conducted under soaked
condition in both one day and one week curing times
according to ASTM D1883-99. CBR tests are conducted witha mechanical loading machine equipped with a movable basethat moved at a uniform rate of 1.27 mm/min. The diameter ofthe piston is 49.6 mm. The loads were recorded as a functionof penetration depth up to a penetration of 10 mm.
4 RESULTS AND DISCUSSION
Bearing ratio tests were carried out for both stabilized andunstabilized clayey soils with different amount of CBR+4 anddifferent curing times at optimum moisture contents values ofeach soils. The stress-penetration curves were plotted andcorrected according to ASTM D1883-99. Fig. 3a shows
typical stress-penetration curves for soil A in one week curing
time and Fig. 3b indicate the same curves for soil B.

(a)

(b)
Fig.3 Typical stress-penetration curves for samples withvarious dose of CBR+4 for one week curing time:(a) Soil A and (b) soil B.

4.1 Effect of Amount of CBR+4 Solution
In this study three different ion exchange solutions doses
were considered for two clayey soils to investigate theinfluence of the amount of this stabilizer on the strength of
soaked samples. CBR values for a penetration depth of
2.5mm were calculated based on ASTM D1883-99 and thefollowing equation:
CBR= [stress at 2.5mm penetration /6.894MPa]100(1)

In the above equation 6.894 MPa is standard stress at 2.5 mm
penetration.
Fig. 4 shows CBR values versus various doses of CBR+4 for
one week curing time. Based on these finding, it can beinferred that dose of CBR+4 doesn’t play any significant rolein the behaviour of soil with higher plasticity index and thiscan be due to less ion exchange. It shows that soil with higher
PI can’t properly accept cation (positive charge) from CBR+4
solution.
Fig. 5 shows the percentage of the increase in CBR values fortwo clayey soils in one week curing time. It shows that withincreasing the dose of CBR+4 up to the 0.12 gr/lit, it increases
CBR values of soil A and this is because of ion exchange
reaction .With increasing CBR+4 more than this value, more
solution without any reaction remain in the soil and this can
301

Fig.4 CBR values versus doses of CBR+4 solution for one
week curing time

Fig.5 Percentage of increase in the CBR values for different
clayey soils

4.2 Effect of the Curing TimeTo investigate the effect of the curing time on the behaviourof the stabilized and unstabilized soil, two curing times wereconsidered; one day and one week.
Fig. 6 illustrates the effect of the CBR+4 on the CBR values
for the various curing times. It shows that the curing times
don’t have any significant effect on the behaviour of soil B
that have higher plasticity index but curing time plays asignificant role in the strength of the soil A with lower PI. It
show that increasing curing time has a significant effect on
the behaviour of soil A and CBR values of soil B don’t show asignificant change. Soil A has a complicated behaviour withadding stabilizer and curing time is an important factor. In
one day curing time, increasing in dose of CBR+4 increase
strength of soil and this is because of higher rate of reaction
with further amount of this solution .With higher curing timeincreasing in dose up to the 0.12 gr/lit increases CBR values
and this is because of ion exchange reaction, but withincreasing CBR+4 more than 0.12 gr/lit, more solution without
any reaction remain in the soil and deteriorate its strength.

(a)

(b)
Fig.6 CBR values versus doses of CBR+4 solution fordifferent clayey soils: (a) Soil A and (b) soil B.

• A greater strength for stabilized soil was achieved forsamples with one week curing time and lower dose of
CBR+4 for soil with lower PI (soil A).
• Increasing in the curing times doesn’t have a constant effecton the strength of the various clayey soils and its effects aredifferent for various soils.
• Adding excessive dose of CBR+4 solution have a negative
effect, because some of this stabilizer remains in the soil
without any reaction and reduces the strength of soil.
• Soil A that has lower PI shows the maximum strength withlower dose of CBR+4 and one week curing time. It shows
168% increase in CBR value.
6 REFERENCES[1] Da Fonseca, A.V., Caberlon Cruz, R., Cesar Consoli, N.
“Strength Properties of Sandy Soil–Cement Admixture,”
Geotech Geol Eng, vol.27,2009, pp. 681–686.
[2] Theng, B.K.G.(1982). “Clay-polymer intractions:
Summary and perspectives,” Clay and Clay Minerals,
vol.30, 1982, pp.1-10.
[3] Dermatas, D., Meng, X.G. “Utilization of fly ash forstabilization /solidification of heavy metal contaminatedsoils,” Engineering Geology, vol.70, 2003, pp. 377-394.
[4] Lahalih, S.M., Ahmed, N. “Effect of new stabilizers on
the compressive strength of dune sand,” Construction
and building Materials, vol.12, 1998, pp. 321-328.
[5] Grim, R.E. “Clay Mineralogy,” Second edition,
McGraw-Hill Book Company, New York, 1968, p. 596.
[6] Nalbantoglu, Z., “Effectiveness of class C fly ash as an
expansive soil stabilizer”. Construction and Building
Materials, vol.18, 2004, pp.377-381.
[7] Nalbantoglu, Z,. Gucbilmez, E.” Improvement of
calcareous expansive soils in semiarid environments,”
Journal of Arid Environments, vol. 47, 2001, pp.
450-463.
[8] CBR PLUS (North America Inc.). “Effective Soil
Stabilization & Dust Control for Roads,” soil
stabilization and dust control.
[9] CBR PLUS (North America Inc.). “Guidelines for the
CON-AID / CBR PLUS Soil Stabilizer,” soil
stabilization and dust control.
[10] CBR PLUS (North America Inc.). “Material Safety Data
Sheet,” soil stabilization and dust control.

5 CONCLUSION
To study the behavior of clayey soils that have been stabilizedwith CBR+4 ion exchange solution, several CBR tests wereconducted on the stabilized and unstabilized soils. The
following conclusion can be drawn from the test results:• CBR+4 solution is a stabilizer that has cation ions and can beused for soils that have anion ions. Two soils that have beenused in this study were clayey soils that have anion ions.
• Plasticity index (PI) plays a fundamental role and this study
shows that the stabilization efficiency decreases withincreasing PI.

302

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Assessment of Low Strain Wave Attenuation for Piles Castin Very Dense Granular Soil
Ramli Nazir, Ph.D, P.Eng. and Osman El Hussien, M.Eng.
Faculty of Civil Engineering, Universiti Teknologi Malaysia

ABSTRACT
In low strain testing of piles, soil within which the pile is embedded causes attenuation of propagated stress wave. The
pile-soil interaction effect on low strain integrity test is not properly identified; however, it may lead to inconclusive
results when piles are embedded in very dense soil. In this paper a method to estimate soil effects on low strain wave
attenuation is proposed. Numerical analysis is utilized to arrive to a theoretical model, which accommodates soil effects
on wave propagation in pile body. The proposed model assuming pile casted in very dense granular soil is comparedwith PIT-S software, which simulates propagation of low strain waves for similar pile-soil conditions. Generally, wave
attenuations estimated by both methods were found to be close for pile subjected to short duration light impact.
Keywords: Pile, Soil, Wave, Attenuation, Low Strain

1. INTRODUCTION
Constructed piles are usually tested in order to evaluate
their integrity as part of foundation construction quality
assurance. The low strain integrity test is routinely
performed, because it is fast and cost effective. This test isbased on wave propagation theory and the name "low strain
test" is based on the fact that when a light impact is applied toa pile it produces a low strain compression wave that travels
down the pile at a constant speed, and generates pile strains ofaround 10-5 micro strain (Hussein and Garlanger, 1992).
Changes in cross sectional area such as a reduction indiameter or material such as a void in concrete produce wave
reflections, which can be used for pile integrity assessment.
This procedure is performed with a hand held hammer togenerate an impact. The accelerometer or geophone is placed
on top of the tested pile to measure the response to thehammer impact via data acquisition and interpretation
through electronic instrument. The hammer impact induces
a one dimensional stress wave into the pile that travels at
speed (c), where (c) is a function of the pile density (ρ) andmaterial elastic modulus (E). The recorded data will beprocessed and interpreted assuming a typical wave speed
value. Pile impedance (Z) is defined as the elastic modulus“E” times the cross sectional area (A) divided by material
wave speed (c) (i.e. Z= EA/c). Impedance changes are
attributed to changes in pile cross sectional area and material
quality. A sample test setup is shown in Figure 1.
The low strain pile top velocity records indicate changes inpile size or quality. However, the pile top motion is also
affected by soil resistance forces. An increase in pile
impedance and/or soil resistance force results in a decrease in
the measured pile top velocity (Johnson and Rausche, 1996).
However, the contribution of surrounding soil on the pile

integrity results is still not precisely quantified. This has led todifficulties in assessment of integrity of piles embedded insoils that produce strong resistance or damping effects on thetested piles.

Figure 1. Components of Pile Integrity Test (PIT) (Massoudi
and Teferra, 2004)
Reflections from defects are also reduced by soil damping
effects. Hence, defects at deeper depths are more difficult to
detect. Rausche et al. (1992) had noted that an impedance
profile analysis, which yields a pile shape as function of depth
must properly identify soil resistance effects before theimpedance profile can be calculated.
While well developed for high strain dynamic tests,
currently there is no generally accepted method available forthe analysis of soil resistance effects on low strain stress wave
propagation. For this reason, engineers have to make certain
assumptions in order to compensate for lack of proper
identification of pile/soil interaction. This may jeopardize thequality of the test output. In addition, soil resistance anddamping effects may lead to inconclusive assessment of pile
integrity, while it is difficult to distinguish the soil response
from the pile response (ASTM, D5882). As a consequence,303

GRL Engineers, the leading firm in the field of pile integrity
testing, has recommended “Inconclusive” test result for piles
subjected to high soil resistance and thus, lacking of clear
stress wave toe reflection (Webster and Rausche, 2011).
2.

METHOD OF ANALYSIS

Assuming a concrete pile shown in Figure 2 having radius(r 0 ), density (ρ c ), modulus of elasticity (E c ), and crosssectional area (A) is subjected to light dynamic force (F 0 )causing displacement (w) in the (z) direction.

A1 = A0 (r0 / r1 ) (γ ) eα ( r0 −r1 )(6)where, (A 0 ), (A 1 ) are vibration amplitudes at (r 0 ) and (r 1 )distances, respectively, (γ) is a coefficient, which depends onwave type and (α) is material damping coefficient.
At the pile toe, the displacement caused by reflected wave
(w r ) in terms of the incident wave caused displacement (w i )is: ρ s vs
1 − ρ v
c c
wr = 
1 + ρ s v s
 ρ v
c c









(7 )

wi

At the pile-soil Interface, the force caused by the reflectedwave (F r ) in terms of the incident wave caused force (F i ) is: ρ s vs ρ v − 1Fr =  c c
 ρ s v s + 1
 ρ c vc


Figure 2. Schematic diagram for the proposed pile and aselected pile segment

where, (ρ s ) is soil density, (v s ) is wave velocity in soil and(v c ) is wave velocity in concrete.
The compressive wave velocity in the concrete pile iscalculated by:Hence, the basic mathematical equations, assuming elastic
soil resistance will be:

Fi − Fi −1 − f i As = ρ c A ∆zwi (t + ∆t ) − wi (t )∆tw − wiFi − f i As = E c A i +1∆z
vi = .

vi (t + ∆t ) − vi (t )
∆t

(8)

Fi

vc =

(1)

Ec

(9)

ρc

The primary wave velocity in the soil is calculated by:

(2)

vs =

(3)

 2 G (1 − ν )  ρ s (1 − 2ν ) (10)where, (v) is particle velocity, (f) is skin friction and (A s ) issurface area.
i.
Sample analysis
The skin friction is assumed to be proportional todisplacement as per the friction-displacement correlation of The following pile data is used in the analysis:
Kraft et al. (1981):
Pile Length, L [m]
=
10
G
=
2.5
Concrete Density, ρ c [t/m3]
(4)
fi =
wi
r
E c [kN/m2]=
4.0E+07
r0 . ln( m )PileRadius,r[m]=0.1784120r0Pile Cross Sectional Area, A [m2] =0.1=
80
where (G) is the soil shear modulus, which can be calculated Number of elements
Pile Element Surface Area, A s [m2] = 0.140124
using equation (5) and (r m ) is the pile influence zone.
Wave Velocity in Concrete, v c [m/s] = 4000
E

G=

s

2(1 +ν )(5)

where (E s ) is soil Young’s modulus and (ν) is soil Poisson’s
ratio.
Since the pile is subjected to dynamic loading, pileinfluence zone can be estimated using ground vibration
attenuation equation (Amick and Gendreau, 2000) as follows:

Very dense granular soil is assumed to surround the pile. Soilparameters are as follows:
Soil Density, ρ s [t/m3]=
Poisson's Ratio, ν=Soil Shear Modulus, G [kN/m2] =

304

1.8
0.4
36000

Wave Velocity in Soil, v s [m/s]Time Interval in Soil, Δt s [sec]γ (equation 6)
α (equation 6)

=
=
=
=

346.41020.000361
1
0.13

Applied dynamic force details are as follows:
Force Amplitude, F 0 [kN]
Pulse duration, [sec]

=
=

1.8
0.00025

The pile influence radial distance (r m ) is calculated using
equation (6) and the abovementioned pile-soil parameters.
Wave amplitude attenuation with radial distance is shown in
Figure 3.
11

Figure 4. Velocity curve at 0.625m depth using PIT-S
0
6
0..6

0..8

0..6

0
4
0..4

Veelooccitty (ccm
V
m/ss)

A
A(rr00)/AA(rr)m=occiitty aatt 05m
25Veello
62
0..6
97
7ccm
m//ss
0..4
49
99
=0
V

0..4

0..2

0
2
0..2

0

0
0
0

2
2

4
4

6
6

nccee, rrR
Raadiaal dissttaan

8
8

1
10
0

--0..2

Figure 3. Wave attenuation with radial distance00

As shown in the figure, the attenuation reaches 95% at 2.6m
radial distance. Hence, r m is considered to be 2.6m. The
proposed model is compared with the output of PIT-S
software manufactured by Pile Dynamic Inc. (PDI). PIT-S
takes a user-input pile shape, soil layer properties andcharacteristics of the low strain hammer impact, and displays
the velocity reflections versus time and versus pile length thata PIT test under such conditions would produce. PIT-S issimulation software in its most basic application, does not
require any data collection. The display of force and velocity
curves or of the velocity curves at two different locations
along the pile (integrity testing of piles integral to a structureis accomplished by analyzing 2 velocity curves) simulate
integrity tests performed with the PIT model FV.
Velocity values at 0.625m, 5.5m ad 9.5m depths,
calculated by PIT-S software and the proposed model are
shown in the following figures:

0
1
0..1

0
2
0..2

T
mss)
mee (m
Tim

0
3
0..3

0
4
0..4

Figure 5. Velocity value at 0.625m depth using the proposed
model

Figure 6. Velocity curve at 5.5m depth using PIT-S

305

0
0..6
6

Table 1. Proposed model and PIT-S produced resultsm=occitty a
5..5
5m
Veelo
att 5
= 0..3720ccm
m/ss

Veelooccitty (ccm
V
m/ss)

0
0..4
4

0
0..2
2

0

--0..2
1..2
2
1

1
1..3
3

1
1..4
4

T
Tim
mee (m
mss)

1
1..5
5

1
1..6
6

Figure 7. Velocity value at 5.5m depth using the proposed
model

Depth
(m)

Time
(ms)

Velocity (Proposed
Model), cm/s

Velocity(PIT-S), cm/s

0.625

0.15625

0.49970.50005.500

1.37500

0.37200.38009.500

2.37500

0.36900.3200The proposed analysis method had been utilized in orderto estimate soil effects on two-way propagated wave after
been reflected at pile toe. Hence, equations (7) and (8) wereused to predict reflected wave characteristics. The pile top
segment force is shown in Figure 10:
44..88meenntt =FN= 44..330066kkN
Foorrccee aatt ttoop sseegm

FN)
orrccee (kNFo

33..88

2..8

1..8

0..8

--0..244..88Figure 8. Velocity curve at 9.5m depth using PIT-S

44..9955

Figure 10. Force curve at pile top segment for the reflected
wave

0
0..6
6

=0
occiitty aatt 9
Veello
36
0..3
V
9..5
5m
m=
69
90
0ccm
m//ss

Hence, the impact force amplitude found to be attenuatedfrom 5.000kN to 4.306kN during its travel downward andupward the pile. This attenuation is caused by soil damping at
toe and skin friction on pile surface. The estimatedattenuation can be summarized as follows:

0
0..4
4

Veelooccitty (ccm
V
m/ss)

T
mss)
mee (m
Tim

0
0..2
2

Table 2: Applied force and encountered attenuations
0--0..2
5
2
2..2
25

2..3
3
2

2
2..3
35
5

2..4
4
2

T
Tim
mee (m
mss)

2
2..4
45
5

2
2..5
5

Figure 9. Velocity value at 9.5m depth using the proposed
model
Summary of velocity values comparison is shown below:

Impact Force amplitude (kN)

5.000

Reflected force amplitude (kN)

4.356

Force amplitude at top pile segment (kN)

4.306

Attenuation at pile toe due to soil damping (%)

11.64

Attenuation due to friction (%)

2.24

Total attenuation (%)

13.88

306

Force amplitude values at 0.625m, 5.5m and 9.5m depths(at 0.15625ms, 1.375ms and 2.375ms time, respectively)
produced from the proposed method of analysis as well as the
PIT-S software is shown in Figure 11.

numerical analysis. The proposed numerical analysis need to
be improved in order to accommodate plastic behavior of soil.
4.

ACKNOWLEDGEMENT

The authors’ wish to thank the Research Management Centreof Universiti Teknologi Malaysia and Pile Dynamic
Incorporation for providing financial support and data,
respectively.
5.

Fig. 11: Force amplitude values using PIT-S and proposed
model
3.

DISCUSSIONS AND CONCLUSIONS

Numerical analysis had been utilized in order to evaluate inorder to evaluate low strain wave attenuation in a concrete
pile subject to soil resistance in a very dense material. The
numerical analysis is based on introducing the skin friction in
the basic numerical equations of motion and Hooke’s law.
The skin friction is assumed to mobilize elastically
proportional to pile displacement. In addition, wave
reflection/attenuation at pile-soil interface is also consideredin the analysis.
Shear stress is assumed to decrease with distance, andconsidered to be negligible beyond a radial distance (pilezone of influence). This radial distance was estimatedassuming that soil does not deform beyond the distance where
vibration amplitude will be attenuated by 95% of its originalvalue.
Short pulse duration was assumed for the light impact
comparable to the pulse chosen in the PIT-S. Despite the factthat PIT-S model includes plastic characteristics of soil,
which is absent in the numerical analysis, the comparison
between the two methods can still be considered as valid dueto the short duration light loading type adopted, which is notexpected to mobilize plastic behavior of soil.
The proposed numerical analysis showed close agreementwith the PIT-S analysis for the selected pile depths (near pile
top, mid of pile and near pile toe). However, the produced
results diverge near pile toe, which can be attributed to thedifferences in evaluation of soil effect at pile-soil interface
between the two methods of analysis.
Low strain F-wave amplitude was found to be attenuatedby 13.88% while traveling 20.0m distance (2-way
propagation) and after been reflected at pile toe, for a selected
10.0 m concrete pile cast in very dense granular soil. Soil
damping at pile toe was found to be responsible of about 84%
of the encountered wave attenuation based on the proposed

REFERENCES[1] American Society of Testing and Materials (ASTM),
2006. Section 4, Vol. 04.09, Standard Test Method for
Low Strain Impact Integrity Testing of Deep
Foundations.
[2] Amick, H., and Gendreau, M., 2000. Construction
Vibrations and Their Impact on Vibration-Sensitive
Facilities. Proc., ASCE Construction Congress 6,
Orlando Florida, pp. 1-10.
[3] Hussein, M. H., Garlanger, J., June 1992. Damage
Detection for Concrete Piles Using a Simple
Nondestructive Method. First International Conference
on Fracture Mechanics of Concrete Structures.
Breckenridge – Colorado, USA.
[4] Johnson, M., Rausche, F., September, 1996. Low Strain
Testing of Piles Utilizing Two Acceleration Signals.
Proceedings of the Fifth International Conference on the
Application of Stress-wave Theory to Piles 1996:
Orlando, USA, FL; pp. 859-869.
[5] Kraft, L. M., Ray, R. P., and Kagawa, T. (1981).
Theoretical t-z curves, Journal Geotechnical Engineering
Division, Proceedings Paper 16653, American Society of
Civil Engineers, Vol 107(GT11).
[6] Massoudi, N., Teferra, W., April, 2004. Non-Destructive
Testing of Piles Using the Low Strain Integrity Method.
Proceedings of the Fifth International Conference on
Case Histories in Geotechnical Engineering: New York,USA, NY. (CD-ROM).
[7] Rausche, F., Likins, G. E., Ren-Kung, S., September,
1992. Pile integrity testing and analysis. Proceedings ofthe Fourth International Conference on the Applicationof Stress-Wave Theory to Piles: The Netherlands; pp.
613-617.
[8] Webster, K., Rausche, F., January, 2011. Pile and Shaft
Integrity Test Results, Classification, Acceptance and/or
Rejection. Compendium of Papers of the Transportation
Research Board (TRB) 90th Annual Meeting:Washington, D.C., USA.

307

Dynamic Soil-Structure Interaction Considering Pore-Water Pressure Coupling
Dr.Qutayba Nazar Al-Saffar*
Civil Engineering Department
University of Mosul / Iraq

Prof. Dr. Akram Y. Al-Sa’aty Prof. Dr. Mohamad T. Al-Layla
Civil Engineering Department
University of Mosul / Iraq

University of Mosul / Iraq

ABSTRACTU

It is a fact that soils are extremely complicated engineering materials
whose constitutive response depends on many factors, including soil type,density, water content, structure, particles size, drainage conditions,
duration of loading, stress history, confining pressure, and stress path,(Corotis et al., 1974).
However, due to the complexity of dynamic soil-structure interaction,
numerical modeling of this phenomenon still remains a challenge. There still
exists many difficulties to cover all the problems listed above in one model.
From the dynamic analysis of soil-structure interaction using non-linear
finiteelement

simulation,

it

has

been

found

that

the

developed

DIANA-SWANDYNE II computer program and the solution algorithms of
the u-p formulation for predicting the coupled behavior of soils are efficient forthe problems solved.
The eight noded isoparametric quadrilateral elements are used in themodeling. Also, a slip-interface element is used for simulation of slippage orseparation at interfaces between soil and structure.
Furthermore, it has been found that the coupling has a significant effecton the response indicated by the lower values of the displacement comparedwith those of uncoupled analysis.
Each of the constitutive models employed in this research proved its
ability to predict the soil response for displacement and excess pore-waterpressure according to the type of material under consideration. But this depends
on the selection and evaluation of the parameters of each constitutive model.

Key words:
Dynamic soil-Structure Interaction, Finite element method, Coupling Phenomena.

* dr.qutaybaalsaffar@yahoo.com or dr.qutayba@gmail.com
308

1. Introduction

soil are usually low. However, systems
subjected to earthquakes or blasts
exhibit large deformations and thus
large strains.
The principal properties that are
required for design purposes are (Das,1983 and El-Sawwaf, 2002):
1. Shear strength and stress strain
characteristics,
2. Liquefaction parameters, such as
cyclic shearing stress ratio, andpore pressure response,
3. Dynamic moduli, such as Young's
modulus and shear modulus,
4. Damping ratio, and
5. Poisson's ratio.
Laboratory measurements aremost widely used for the determinationof dynamic properties and cyclic
strength of soils, for their accuracy andability to cover various strain ranges(Ishihara, 1996).
The soil deposits, which are
subjected to earthquake loadings arealso capable of these two behaviors,
i.e. flow liquefaction and cyclic
mobility (Casagrande, 1975). Thesebehaviors are best considered in termsofsteady-statesoilmechanics(Casagrande, 1975; Castro andPoulos, 1977).
The finite element techniques
are now the most widely used tosimulate the dynamic soil behavior
because of the ease and realismwith which the geometry andmaterial properties can be modeled(Al-Saffar, 2009 ).

Soil mechanics is known as thescience that deals with the properties
and behavior of soils under stresses.
However, only static stresses are mostoften implied. On the other hand, soil
dynamics is that branch of soil
mechanics
which
deals
with
engineering properties and behavior ofsoil under the variation of stresses
imposedbydynamic
loads(Das, 1983).
The behavior of soils underdynamic loading deals with additional
problems due to diversity in the areasand complexity in the conditions underwhich dynamic loads are applied(Prakash, 1981)
Ishihara (1967) derived theequations of motion for a saturated
porous elastic matrix for the twoconstituents of the soil separately.
He related the four elastic constants of
Biot's theory to the compressibility ofthe bulk soil, the solid material and thewater. He concluded that the waves
travelthroughthesolid-water
system without causing any change inthe pore volume.
Bazant and Krizek (1975)
extended Biot's linear elastic theory to
the non-linear non-elastic case. They
formulated an incremental stress-strain
relationship that takes into account the
non-linearity and non-elasticity of thesoil.
Zienkiewicz and Bettess (1982)
extended Biot's formulation to rocks
and other saturated porous materials.
They also presented a physicalderivation of the governing equations.
The behavior of soil isnonlinear and the dynamic properties
are substantially dependent on thestrains caused by the dynamic loads
applied to the soil. For instance, thedynamic loads due to machinefoundation, the amplitude of motion
and, consequently, the strains in the

1. soil-structure interaction

A large part of geotechnicalengineering is directly concerned withsoil-structure interaction. In somecases, the soil may be modeled in great
detailandwithcomplicatedconstitutive relations while thestructure is represented very simply. In
dynamic problems, the addition of time
as a variable increases the mass of data
and the number of options.
309

There is no universal, ideal, orcomplete way to deal with problems ofsoil-structure interaction. The choicemust be based on the relative
importance of the various parts of the
problems and on the time andresources available to deal with it(Salih, 2001). Dynamic interaction
between soil and structure is defined asa phenomenon of transmitting kinetic
energy through the interface media.
The use of multistory buildings
with basement floors is very importantin overcrowded cities. Considering the
dynamic behavior of these types ofstructures under earthquake loading,many researches have neglected the
dynamic interaction between soil,
substructure and superstructure. They
merely simulate the soil using a-seriesof springs and viscous damping locatedat the base of the structure only or use
the shear beam analogy method.
There are several factors whichmust be considered in the design ofhigh-rise buildings. Some of thesefactors are related to the superstructuresuch as strength, serviceability andductility while others are soil related,such as settlement.
Either continuum approach or
finite element approach can solve
soil-structure interaction problem.

the coupling occurs through thegoverning
differential
equationsdescribing
different
physicalphenomena. The main example of thistype of coupling is the soil-pore waterinteraction.
It is well known that the
behavior of soil is strongly influenced
by the pressure of the fluid(generally water) present in the poresof the material, (Zienkiewicz &Taylor, 2000).

3. Constitutive relations
The concept of the effective
stress is here of paramount importance,thus if σ describes the total stress(positive in tension) acting on the total
area of the soil and pores, and p is thepressure of the water (positive incompression) in the pores, the effective
stress is defined as:

σ ′ = σ + mp
where:

mT = [1,1,1, 0, 0, 0]---(1)---(2)

Assuming that the soil can be
represented by a linear elastic model
we have:σ ′ = Dε2. Coupling Phenomena---(3)

Immediately the total discrete
equilibrium equations for the solidfluid mixture can be written in exactly
the same form as is done for all
problems of solid mechanics:

Coupled systems are those
applicable to multiple field anddependent variables, which interact
with each other, where the solution of
any system being impossible without
simultaneous solution of the other.
It is convenient to classify
coupled systems into two categories :Class I Problems : This class contains
problems in which coupling occurs on
domain interfaces via the boundary
conditions imposed there.
Class II Problems : This class contains
problems in which the various domains
overlap either totally or partially. Here,Mu + Cu + ∫ B T σ dΩ
Ω

+ f =

0

---(4)
Where:
M, C and f have the usual meaning of
mass, damping and force matrices,
respectively.
B : is the strain-displacement matrix .
Ω : is the numerical integration
domain.
310u : is the displacement matrix in the

We shall now consider the coupledproblem discretized in the standard
manner with the displacement vector
approximated as:

global direction.
The standard stiffness matrix Ku ≈ u = Nuu---(5)and the fluid similarly approximatedin equation (8) can be represented as:



T

Ω



p ≈ p = Np p
where


u and


p

are

the

Ω

and the coupling matrix Q can be

nodal

represented as:

Q = ∫Ω B mN p dΩ
TNow, the term involving the---(10)

Such damping matrices C havestress must be split as:a physical significance and are always
introduced in earthquake analyses or∫ B σ dΩ = ∫ B σ ′ dΩ − ∫ B mp dΩTΩ---(9)

N p are appropriate shape function.

TT---(6)

parameters of each field and Nu andΩ∫ B σ ′dΩ =  ∫ B DB dΩ u = Ku

as:TsimilarΩ---(7)

problemsofstructuraldynamics. With the lack of any special

To allow the direct relationshipinformation

between effective stresses and strain

about

the

natureofdamping, it is usual to assume the so(and hence displacements) to becalled 'Rayleigh damping' in which:incorporated. For a linear elastic soil

C = αM + β K---(11), where α and β are coefficientsskeleton we immediately have:Mu + Cu + Ku − Q p + f = 0

determined by experience (see for
instance Clough and Penzien, 1975 ;---(8)

Seed et al., 1975).

Equation (8) represents theIndynamic equation of motion for the

the

above

conventional

discretization, the same element shapes

solid and fluid phase of the saturated

are used for the u and p variables,

porous media subjected to cyclic

thoughloading, where:notinterpolations.

M : is the global mass matrix.

necessarily

identical

With

dynamic

theequations coupled to the pressure field,

K : is the global stiffness matrix.

an

C : is the global damping matrix.

additional

equationis clearlyneeded from the transient seepage

Q : is the coupling matrix.

equation of the form:p : is the global pore pressure matrix.

− ∇T (k∇p ) +f : is the forces matrix for the solid
phase.
3111p + εv = 0 ---(12)QQ is related to thewherecompressibility of the fluid , k is the

4. DIANA-SWAN DYNE II Program

DIANA-SWANDYNE II or
Dynamic Interaction and Nonlinear
Analysis-SWANseaDYNamic program
version II is an improved version of the
DIANA-SWANDYNE I
The program is developed by
Prof. A.H.C. Chan based on his thesis(Chan 1988) under the supervision ofProf. O.C. Zienkiewciz.
The program is intended for
Static, Consolidation and Dynamic
analysisforproblems
inGeomechanics. It is a 2-Dimensional
program which incorporates Planestrain and Axisymmetric analysis. The
Governing equation that is being
solved is the Fully Coupled, Large
Deformation Biot Equation with theu-p simplification with the fluid
acceleration neglected.
DIANA-SWANDYNE II iscapable of performing analysis forstatic (drained and undrained).
Consolidation and dynamic (drained
and undrained). The solid and fluid aretreated as separate phase in the
program and different integration rulecan be used for each of them.
The program is based on finite
element method with triangular andquadrilateral isoperimetric elements inspatial domain. The time integration isdone with the Generalized Newmark
method (Newrmark 1959).

permeability and εv is the volumetricstrain in the soil skeleton , which on
discretization of displacement is givenby:
ε v = mT ε = mT Bu---(13)The equation of seepage can
now be discretized in the standard
Galerkin manner as:







---(14)

QT u + Sp + Hp + q = 0

Equation (14) represents the dynamic
equation of motion for the fluid phaseofthesaturatedporousmedia

subjected to cyclic loading, where:
S : is the compressibility matrix.
H : is the permeability matrix.
q : is the forces vector for the fluid
phase.
The standard compressibility
matrix (S) and permeability matrix (H)in equation (14) can be represented asfollows:
S = ∫ N TpΩ1
N p dΩ , andQ

H = ∫ (∇N p )T k∇N p dΩΩ

5. Verifications of the program---(15)

Finaly, from the two main dynamic
equations of motion (8 and 14) , theformat of the coupled equations using
u-p formlation can be written as :
M
0



0  u   C  +  T0  p  Q---(15)
0   u   K
   + S   p  0f− Q  u    = −  H   p 
q 

312

In order to examine theapproximations introduced in the (u-p)
form and to verify the performance ofthe developed DIANA-SWANDYNE
II program, problems of onedimensional soil layer is adopted andimplemented in the numerical solution
procedure for this purpose.

Numerical Application
Consolidation Problems
Two problems are chosen to besolved using DIANA-SWANDYNE II,
in the first problem a simulation of
consolidation test is performed while,the second problem is conducted tosimulate the long term field settlement.
A: Consolidation of Swansea Blue
Clay in a Rowe Consolidation Cell
Lewis and Schrefler (2000)
simulated the experimental results ofRowe consolidation cell using finite
element program and a good
agreement results are obtained. In thiswork the simulation of this test are
performed and a comparison of theresults are discussed.
Problem definition :The problem represents anexperimental
consolidation
test
performed on a soil sample of Swansea
blue Clay in (10") Rowe consolidation
cell. Large sample of 10" (254 mm) indiameter and (61 mm) in height istested (figure 1-a). One stage ofconsolidation loads are simulated whena load of (172 kPa) is applied and thesoil properties are shown in table (1).
The finite element discretization isshown in figure (1-b), where 130
elements are used to simulate the test.
The solid skeleton and the pore waterin the soil are each modeled with 8nodded and 4-nodded serendipity
isoparametric elements respectively. At
the vertical boundaries of the solid
skeleton, only vertical movement ispermitted (i.e. vertical rollers). The
water level is assumed to be at the
ground surface (i.e. saturated state) andthe pore water pressures at the freesurface are taken as zero. A non-linear
elasto-plastic model is used in finite
element simulation of the soil
behavior.
Numerical Analysis Results:
Figure
(2-a)
show
the
settlement behavior of the middle
surfaceof the sample. The results
obtained by the present study using
DIANA-SWANDYNE II program
found to be acceptable values with theresults given by Lewis and Schrefler(2000) using non-linear consolidation
finite element program. The responseof the pore water pressure with time

during the test are shown in figure(2-b) and the comparison of the results
indicates that the numerical resultsfrom the present study are in good
agreement with the numerical results
given by Lewis and Schrefler (2000).
The small difference in results is due todifferent models used in simulation.
B: Two-Dimensional Consolidation• Problem definition :A two dimensional plain strain
consolidation problem is solved in thissection. The problem is similar to thatsolved by Siriwardane and Desia(1981)* using elastic and elasto-plastic
models, also the same problem issolved by Lewis and Schrefler (2000)
using non-linear consolidation finite
element program.
The
adoptedfinite element mesh are shown infigure (3) where 30 elements have beenused where the solid skeleton and thepore water in the soil are each modeledwith
8-noddedand4-nodded
serendipity isoparametric plane strain
elements respectively. A Cam-Claymodel (critical-state model type) isused in the present study. The
properties of the soil and parameters ofthe model used in simulation are
shown in table (2).
An external load applied at the
top surface of the model for a length
B= 10 varies with time as indicated infigure (3).
Numerical Analysis Results:The
predicted
pore-waterpressure using Cam-Clay model arecompared with the results of
Siriwardane and Desia (1981) and theresults of Lewis and Schrefler (2000)
in figure (4). The simulated pore-waterpressure are non-dimensional withrespect to initial applied load
(po= 1000), also, the non-dimensionaltime factor defined as (Tv = cvt/H2).
A good agreement of results
can be observed in the comparison andthe small difference between models'
results are due to the inclusion of theplasticity effects which decreases themagnitude of dissipation of pore waterpressure for almost the whole time
span. Therefore, the predicted porewater pressure is higher than thosefrom an elastic analysis.
313

Dynamic
Analysisof
SoilFoundation Interaction
In this problem, the dynamic
analysisoftwo-dimensionalsoil-foundation system is considered
and the response of the system to thedifferent type of dynamic loading arepresented. In addition, the effect ofpore-water pressure coupling and theinterface-element are considered.
Problem definition :Figure
(5-a)
shows
a
(2.0*0.25m) concrete strip footing
supported on (H=20.0m) saturated soil
stratum laying over a rigid bedrock.
The soil properties are shown in table(3) while, the properties of the footing
material (concrete) are shown in table(4). The footing is subjected to surface
excitations (step, ramp and Sinusoidal
function), which are the same as thosegiven in (table 1). The same problem
was solved and discussed by Abbas(2003) except that he used highfrequency
sinusoidal
loading( ω = 66.92 rad/sec), and in the present
study, an elastic and elasto-plastic
model is used to solved this problem
using
DIANA-SWANDYNE
II
program.
The selected depth of thesaturated layer of (20.0m) is taken toassure that the ratio (h/b= 20) is wellgreater than the minimum indicated by
Japon et.al (1997)*, to minimize theeffect of the rigid bedrock on the
dynamic response of the footing.
The
finiteelement
discretization is shown in figure (5-b),
where
22 elements are used tosimulate the soil-foundation system.
The solid skeleton and the pore waterin the soil are each modeled with 8nodded and 4-nodded serendipity
isoparametric plane strain elements
respectively. At the vertical boundariesof the solid skeleton, only verticalmovements are permitted (i.e. verticalrollers). The water level is assumed to
be at the ground surface (i.e. saturated
state) and the pore water pressures at
the free surface are taken as zero. A
non-linear elasto-plastic model is usedin finite element simulation of the soil
behavior.

Numerical Analysis Results:
Figure (6) shows the responseof the vertical displacement at
point
A for the strip footing when it issubjected to step and ramp surface
excitations. The upper part of thisfigure present the result of Abbas(2003) while, the lower part shows theresults of the present study.
Furthermore, in the right side of thetwo parts the coupling effect is present.
The comparison of the results indicatesthat the results of the present study arein good agreement with the results
given by Abbas (2003).
Figures (7) shows the responseof the vertical displacement at point A
for the strip footing when it issubjected to surface excitations of highfrequency
sinusoidal
loading( ω = 66.92 rad/sec) for both dry andsaturated state , and the coupling effectis clearly shown in figure (8).
The
excesspore-waterpressure-time history at point B for the
strip footing when it subjected tosinusoidal excitation is shown in figure(9).
In all of the results above, norelative motions allowed at theinterface between soil and foundation(without interface element). Asimple
slip-interface model is used in thepresent study which is describedbriefly in chapter three with material
properties same as the soil properties.
The effects of using this model on thedisplacement and excess pore-waterpressure response are shown in figures(10 and 11) respectively, and theinterface effect is clearly shown infigures (12 and 13) respectively.
The comparison of the results
indicates that the results of the present
study are in good agreement with theresults given by Abbas (2003). The
differences in results are due todifferent soil models and different
interface models used in simulation.
•314

9. References

8. Conclusions

From the studying of dynamic
analysis of soil-structure interaction
using non-linear finite element
simulation,
thefollowingareconcluded:
1. Thedeveloped
DIANASWANDYNE II computer program
and the solution algorithms of theu-p formulation for predicting thecoupled behavior of soils are foundto be efficient for the problems
solved.
2. The finite element method
represents a powerful tool in soil
dynamic in which there are no
analytical solutions available yet.
3. Two-phase materials (such as
saturated soil) subjected to
dynamic
loading
couldbeformulated
withapproximate
numerical solutions with anacceptable degree of accuracy.
4. The coupling has a significant
effect on the response indicated by
the
lower
valuesofthedisplacement compared with thoseof uncoupled analysis. This is dueto the role of water in the pores
which shares the solid in carrying
the applied load at the earlier
stages. Then, diminishes as theexcess pore water pressure isdissipated with the progress oftime.
5. Each of the constitutive models
employed in this study proved its
ability to predict the soil response
in displacement and excess porewater pressure according to thetypeofmaterial
underconsideration. But this, certainly,
depends on the selection and theevaluation of the parameters ofeach constitutive model.

1. Al-Saffar, Q.N. (2009),"Simulationof Soil Dynamic Properties andPore-Water Pressure Coupling
Using Finite Element Method",Ph.D.
Thesis,CollegeofEngineering, University of Mosul,
Iraq.
2. Bazant, Z.P. and Krizek R. J.,(1975), "Saturated Sand as an
Inelastic Two-Phase Medium".
Journal of Engineering Mechanics,
ASCE, Volume 101, No. EM4,
pp.317-332.
3. Bowles, J.E., (1996), "Foundation
Analysis and Design", Mc-GrawHill Book Co., New York.
4. Casagrande,A.,(1975)"Liquefaction And Cyclic Mobility
Of Sands. A Critical Review"
Proceedings of the 5th Pan
American Conference on Soil
Mechanics
andFoundation
Engineering, Buenas Aires, Vol. 5,
pp. 80-133.
5. Castro, G. and Poulos, S.J., (1977)"Factors Affecting Liquefaction
And Cyclic Mobility"Journal of the
GeotechnicalEngineering
Division, ASCE, Vol. 103(6),pp.501 - 516.
6. Chopra A.K., (1995), "Dynamics ofStructures-Theory and Applications
toEarthquake
Engineering",
Prentice-Hall, Inc., New Jersey,USA.
7. Das, B.M., (1983), "Fundamentals
of Soil Dynamics", Elsevier
Science Publishing, New York.
8. Dinesh, S.V., Sitharam, T,G. and
Vinod, J.S., (2004), "Dynamic
Properties
And
Liquefaction
Behavior of Granular Materials
Using Discrete Element Method",
Current Science- Geotechnics
And Earthquake Hazards-, Vol. 87,
No. 10.
9. El-Sawwaf, M.A.M., (2002), "The
Effect of Dynamic Loading on The
Behavior of Shallow Foundation",
Ph.D.Thesis, University of Cairo,
Egypt.
315

Townsend, F.C., Valeara, J.E., andWilson, J.H., (1976), "Cyclic
Triaxial Strength of Standard Test
Sand", Journal of the GeotechnicalEngineering Division, ASCE, Vol.
102,GT5, pp.511 - 523.

10. Ishihara, K., (1967), "Propagationof Compressional Waves in a
Saturated Soil". Proceedings ofInternational Conference on Wave
Propagation
andDynamic
Properties of Earth Materials,
University of New Mexico,
Albuquerque, New Mexico, August23-25, pp. 451-467.

16. Srbulov, M., (2008)", Geotechnical
Earthquake
EngineeringSimplified Analyses with Case
Studies and Examples", Springer
Science and Business Media B.V.,
United Kingdom .

11. Ishihara, K., (1996), "Soil Behavior
inEarthquake Geotechnics",Oxford
University Press Inc.,New York,USA.

17. Yang, Z., Elgamal, A. and Parra,E., (2003), "Computational Model
for Cyclic Mobility and AssociatedShear Deformation", Jornal of
Geotechnical and Geoenviromental
Engineering, pp. 1119-1127.

12. Khosla, V., (1985), "Advances in
The Art of Testing Soils Under
Cyclic Condations", The American
Society of Civil Engineering, USA.
13. Polito, C.P., (1999), "The Effects
Of Non-Plastic and Plastic Fines
On The Liquefaction Of Sandy
Soils", Ph.D. Thesis, Virginia
Polytechnic Institute and State
University.
14. Prakash,
S.,(1981),
Dynamics", McGraw-Hill
Company, New York.

18. Zienkiewicz, O.C., Leung, K.H.,
Hinton, E. and Chang, C. T.,(1982),"Liquefaction
andPermanent Deformation under
Dynamic
Conditions-Numerical
SolutionandConstitutive
Relations", Chapter 5 in Soil
Mechanics-Transient and Cyclic
Loads, (edited by G. N. Pande andO.C. Zienkiewicz), pp. 71-103.

"Soil
Book

15. Silver, M.L., Chan, C.K., Ladd,
R.S., Lee, K.L., Tiedmann, D.A.,
Table (1):Table (2):Properties of the Swansea Blue Clay

Properties and parameters used in simulation

13000

Elastic modulus

Ei = 26000.0 kPa

Elastic modulus (Ei)Poisson's ratio

υ = 0.20

Poisson's ratio ( υ )0.4

Solid skeleton density

ρ s = 18.0 kN/m

Soil density( γs)

18.5

Water density ( γw)

10.0

Void ratio (eo)0.9Porosity (n)0.473

cm/minPermeability (kx=ky)

4.0*10-5

φ ′ = 8.0 o

M

1.05

c ′ = 22.8 kPa

λ
κ

0.14
0.05

*cv

0.282

Water density
Permeability
Angle of internal
friction

Effective cohesion3

ρ f = 10.0 kN/m3
K = 4.83*10-6*Calculated from Tv = cvt/H2
316

Table (3):Table (4):Prosperities for the Saturated Soil Layer

Elastic Concrete Prosperities

Elastic modulus

E = 30000.0 kPa

Elastic modulus

Poisson's ratio

υ = 0.2

E = 42 * 10 6 kPa

Poisson's ratio

υ = 0.2

Solid weight density*ρ s = 30.01 kN/m

Weight density

ρ = 23.2 kN/m3

Water weight densityρ f = 10.0 kN/m3

Permeability

K = 0.0015 m/sec

Porosityn = 0.333

Solid bulk modulus

Ks = ∞

Pore water bulk modulus

K f = 400000.0 kPa

3*Solid mass density = 3.001 gm/cm3= 30.01 kN/m3(weight density )254 x 61 mm soil sample

a- Dimension of Sample in Rowe Cell
Surface Traction = 0.172 MPa
PWP = 0

H= 61 mm

D= 10" =254 mm

b- Finite Element Idealization
Figure (1) Finite Element Simulation for Rowe Consolidation Cell

317

Time (min)Pore Water Pressure (kPa)

Settelement (mm)

0

10

20

30

40

0
0.05
0.1
0.15
0.2
0.25
0.3
0.35
0.4
0.45
0.5

50

60

70

80

90

100

Lewis and Schrefler (2000)

Present study

-a-

200

Lewis and Schrefler (2000)

175

Present study

150
125
100
75
50
25

-b-

0
0

10

20

30

40

50

60

70

80

90

100

Time (min)Load, Po

Figure (2): Results of the Simulation of Rowe Consolidation Cella- Settlement versus time for clayey soil.
b- Pore-water pressure versus time for clayey soil.

0.07

B

Time factor, Tv

2B

6B

Figure (3): Finite Element Idealization and Load History
318

Elastic analysis:
Siriwardane and Desia (1981)++ Lewis and Schrefler (2000)
Elasto-plastic analysis:
Siriwardane and Desia (1981)
oo Lewis and Schrefler (2000)

a- Results of: Siriwardane and Desia (1981); Lewis and Schrefler (2000)

1.00

Elastic
Cam-Clay

Z/H

0.75
0.50

381 days0.25

181 days

21 days

0.00
0

5

10

15

20

25

p x100/poxb- Results of the present study
Figure (4): Pore-water Pressure Versus Depth Along Saction A-A
at Three Different Time Values:(1) Tv= 0.059, (2) Tv= 0.504 and (3) Tv= 1.060

319

30

2x0.25 m concrete
strip footing

W.T.

Free drainage
surface
Saturated
Soil Stratum

Interface

20 m
Impermeable
bedrockA(a): Soil-Footing System

Free drainage
surface

Impermeable andSmooth Boundary

B

Fixed and
Impermeable
Boundary

40 m(b): Finite element idealization
Figure (5): Strip Footing on Saturated Soil Stratum

320

0.8

0.8

1

Step
Ramp

0.7

0.6

Displacement (mm)Displacement (mm)

0.6
0.5
0.4
0.3

0.5

0.3
0.2

0.1

0.1

0

0

0.025

Results of(Abbas, 2003)

0.4

0.2

0

0

0.05

0.025

0.05

Time (Sec)Time (Sec)
0.8

0.8

1

Step
Ramp

0.7

2

Step
Ramp

0.7

0.6

0.6

Displacemebt (mm)Displacement (mm)

2

Step
Ramp

0.7

0.5
0.4
0.3

0.5

0.3

0.2

0.2

0.1

0.1

0

0

0

0.025

0.05

Time (Sec)Present Study
Results

0.4

0

0.025

0.05

Time (Sec)Figure (6): Vertical Displacement Time History at Point A for the Strip Footing
1: without Coupling (Dry State), 2: with Coupling Effect (Saturation State)

321

0.5

Results of Abbas (2003)

Displacement (mm)Present study results
0.25

0

-0.25

-a-

-0.5
0

0.2

0.4

0.6

0.8

1

1.2

Time (Sec)0.4

Results of Abbas (2003)

Displacement (mm)Present study results
0.2

0

-0.2

-b-

-0.4
0

0.2

0.4

0.6

0.8

1

1.2

Time (Sec)Figure (7): Vertical Displacement Time History at Point A for the Strip Footing
a- without Coupling (Dry State), b- with Coupling Effect (Saturation State)322

Displacement (mm)

0.5

0.25

0

-0.25

Without coupling effectWith coupling effect

-0.5
0

0.2

0.4

0.6

0.8

1

1.2

Time (Sec)

Excess pore-water pressure (kPa)

Figure (8): Coupling Effects on Vertical Displacement for theStrip Footing (present study results)
1.5

Results of Abbas (2003)

Present study results

1.2
0.9
0.6
0.3
0
-0.3
-0.6
-0.9
-1.2
-1.5
0

0.2

0.4

0.6

0.8

1

Time (Sec)Figure (9): Excess Pore-water Pressure Time History at Point B for Strip
Footing Subjected to Sinusoidal Excitation

323

0.4

Displacement (mm)

0.3
0.2
0.1
0
-0.1
-0.2

Results of Abbas (2003)

Present study results

-0.3
0

0.2

0.4

0.6

0.8

1

Time (Sec)

Excess pore-water pressure (kPa)

Figure (10): Interface Effects on Vertical Displacement
for the Strip Footing

1.6
1.2
0.8
0.4
0
-0.4
-0.8
-1.2

Results of Abbas (2003)

Present study results

-1.6
0

0.2

0.4

0.6

0.8

Time (Sec)Figure (11): Interface Effects on Excess Pore-water Pressure for the Strip
Footing

324

1

0.4

Displacement (mm)

0.3
0.2
0.1
0
-0.1
-0.2
-0.3

Without interface

With interface

-0.4
0

0.2

0.4

0.6

0.8

1

Time (Sec)

Excess pore-water pressure (kPa)

Figure (12): Interface Effects on Vertical Displacement for the Strip
Footing (present study results)

1.6
1.2
0.8
0.4
0
-0.4
-0.8
-1.2

Without interface

With interface

-1.6
0

0.2

0.4

0.6

0.8

1

Time (Sec)Figure (13): Interface Effects on Excess Pore-water Pressure for the Strip
Footing (present study results)

325

Analysis And Design Of Tunnel Lining In Alluvial Soil Deposits

ABSTRACTProf. Dr. Nuri al-Mohamadi
Applied Science University ,Amman , Jordan ,11931 .
Formerly ; Consultant for Baghdad Metro Project ; Baghdad Rapid Transit Authority
e-mail: nalmohamadi@yahoo.com

This paper is concerned with analysis and design of a typical tunnel section of the proposed Baghdad Metro Project. A crosssection of the soil profile is presented showing soil layering, which consist of a top cohesive layer underlained by a lower sandlayer of fine to medium sand. A summary of soil characteristics required for tunnel analysis is given. Two loading cases are
considered for the analysis of the driven bored tunnel. Load case 1 ( LC 1 ) the water table is assumed at ground surface , whileload case 2 ( LC 2 ) the water table is assumed at bottom of tunnel. The earth pressures and water pressures acting on thetunnel for these loading cases are calculated . The tunnel lining consists of seven precast segments of 0.3 m thickness . Twotunnels of diameter D= 6.2m are driven with a distance= 16.5m between center of tunnels . This distance becomes 9.0m when
the tunnels approach to a station. A finite element program is used to analyze the normal forces, bending moments, shear
forces and displacement of the tunnel lining . The tunnel section is idealized as a framework consisting of (a) straight beams
representing the reinforced concrete segments, (b) springs which represent the soil .Spring stiffness factor is calculated fromelastic theory and soil characteristics. The maximum values of normal forces, shear forces bending moments anddisplacements are presented.
Keywords: Alluvial soil, loading cases, precast concrete segments, tunnel lining, structural analysis.
1. INTRODUCTION
The development of transportation in large cities and thecontinuous increase of population require the constructionof underground transportation systems.
The relative positions of tunnels and the construction
procedure affect the soil deformation and internal forces
shear forces and bending moments in the lining , [4].
Precast segmental linings are used in circular tunnels thatare bored using shield tunnel boring machines, mainly insoft ground. shield tunnel lining in soft ground is usuallydesigned based on empirical analytical methods with some
simplifying assumptions related to soil homogeneity,
stiffness and height above the tunnel, [8]. Several curved
precast reinforced concrete segments are assembled inside
the tunnel at the tail of the boring machine to from acomplete ring. The number of segments is a function oftunnel diameter and construction method. The segments
are supplied with water proofing neoprene gaskets on eachedge and bolted together with high strength grip bolts tocompress the gaskets and assure impermeability of thelining.
2. SOIL CONDUCTION AND GEOTECHNICALASSESMENTS

Figure 1: Soil profile and elevation of metro rail level
The moisture of the upper cohesive soils, usually in therange 20% to 30%, which is close to and often lower thanthe plastic limit. This conforms with the desiccated natureof the soils due to high evaporation during precipitation
associated with high shrinkage stresses. The resulting soil
characteristics
are;
over
consolidation,
reducedcompressibility and high strength. Therefore these soils aredescribed as stiff or even hard than would be expectedfrom normally consolidated alluvial deposits in moretemperate areas, [7], [2].the presence of sand stratum
below the main cohesive layer and the highly variable soil
conditions with all their random lensing, in the presence ofhigh water table give rise to some of the more difficult
problems of constructing deep underground structures,
especially in the urban area where adjacent buildings may
be affected and services and access have to be maintained.
Tunneling will always be below water table either in thesilty fine sands or more often through a silty clay – silty
sand interface. All excavations must be supported withgood excavation support system with minimum lateralyield to protect adjacent buildings. Control of ground

In Baghdad region the flood plane alluvial deposits extend
to a large depth below the investigated depths of 40m and
50m. in general the top layers are cohesive layers of silty
clays or clayey silts mostly of intermediate to highplasticity, overlying medium to dense or very dense silty
fine sands. The upper cohesive soils contain lenses of silty
fine sands, are in the order of 10 to 20m deep, occasionally
may be only 3m deep and exceptionally more than 35m
deep. The silty sands contain numerous randomly
occurring inter-layers and lenses of cohesive soils ofvarying thickness and extent as well as pockets or lenses ofgravelly sands. Groundwater levels are in general between
1m and 3m deep; these details are shown in figure 1.

326

water will be essential. Recharge in nearby areas may berequired to protect nearby buildings from excessive
settlement and differential settlement due to water table
lowering for prolonged periods.

assumed to take into consideration the effect of adjacentbuildings and street loads for both loading cases.
4.2 Load case 2 (LC2)This load case simulates the loading of excavated tunnel
adjacent to each station from both sides and extends toabout 10 to 15 m from the station. The water is assumed atbottom of the tunnel, figure 2.

3. GEOTECHNICAL ASSESMENT
The main geotechnical parameters required to the designof segmental tunnel lining and excavated stations are
summarized in table 1,[7].
Table (1a&b): Summary of Geotechnical Parameters(a) cohesive soils
Parameters
Value
Bulk density (γ b )19.1 KN/m3(1.95 t/m3)
Shear Strength
Undrained (C u )75 KN/m2
Drained (Φ’)
25°
Elastic Modulus (E)15 MN/m2 = 200 C u
Poisson’s Ratio (ν)
0.5
Consolidation characteristics:
Compression Index (C c )0.2 for depth < 13m
0.3 for depth > 13m
Swelling or rebound Index(C r )
Initial Void Ratio (e o )Mainpre-consolidationpressure (P c ’)Earth Pressure at Rest (K o )Sulphate Content (SO 3 )(b) Silty Sands
Parameter
Bulk Density (γ b )RR0.05 (0-15m) depth
0.06 (15-22m) depth
0.65 (0-13m)
0.72 (13-22m)
325 KN/m2

R

R

The distance between the two tunnels is minimum (d 2 =9.0m), so the effect of the adjacent second tunnel
excavation, should be considered, i.e. the soil will be loose
and the elastic modulus (compression modulus of springs)
must be reduced around the horizontal center line of thetunnel facing the first tunnel, from 30 MN/m2 to 20
MN/m2. The calculated earth pressures and water pressures
acting on the tunnel for load cases LC 1 & LC 2 are given
in figures 3a & 3b.
R

1.2 at 3m depth
0.7 at 25m depth
Range 0.1 to 8.7 g/l

P

P

R

P

P

Value
19.1 KN/m3(1.95 t/m3)
35°
P

P

Effective Angle Of Shearing
Resistance (Φ')
Elastic Modulus (Es)
Poisson's Ratio(ν)
Earth Pressure Coefficient at
Rest (K o )Permeability(K)Figure 2: Loading cases LC1, LC2

P

30 to 50 MN/m2
0.3
0.7 to 8m depth
0.5 at 25 depth
Order of 10-6 m/s
PPP

4. LOAD CASES AND GROUND WATER
Tow load cases are considered ;4.1 load case1 (LC1)
With highest soil cover above tunnel crown (~10m) andthe distance between the two tunnels is large (d ≈ 16.5m).
The ground water table is assumed at ground surface,
figure2. in this loading case the tunnel lining will be
subjected to the greatest normal force, so this loading casewill be decisive for evaluating the normal stresses in the
joints of the tunnel segments. Since the distance between
the centerline of the two tunnels is wide enough (d ≈16.5m), the influence of the adjacent second tunnel driving
or excavation is of minor importance and it will not benecessary to be considered for the calculation. Thisloading case simulates the actual loading for bored tunnel.
A surcharge load equivalent to a soil cover of 1.5m is

Figure 3a: Earth pressures and water pressures acting ontunnel, LC 1

327

6.CALCULATION OF SPRING CONSTANT(MODULUS OF ELASTIC SUPPORT)The elastic deformation of the tunnel can be simulated bymeans of springs situated normal to the ring at constantsmall distances about (0.6m) so that the soil characteristics
are represented as close as possible to the actual case, as
shown in figures 4&5.

Figure 3b: Earth pressures acting on tunnel, LC 2
5.DETERMINATION OF COMPRESSION MODULUS E s
The compression modulus E S is required for the finite element
analysis of tunnel lining. E S can be obtained from consolidation
tests results presented as void ratio e vs. log applied effective
stress σ v ’.

∆σ V∆e∆e = eo − (1 − eo ) ∗ e
ES =Figure 4: Simulation of segmental lining model

′

(1)
(2)

Where
∆σ v ’= change of applied effective stress.
∆e = change of void ratio.
e o = initial void ratio.
e = void ratio corresponding to any stress.
Since most tunneling lies in a mixed soil of clay, silt andsand, so an average value of E s for different soil types.
The consolidation test results are not presented due tospace limitations.
Table (2): Calculation of compression modulus E s
σ v’
Soil
eo
e
∆e
∆σ v ’
KN/m2
type
KN/m2
Clay 0.52 50
0.486 0.287 1550
1600
0.420 0.318
Silt
0.6
50
0.580 0.368 750
800
0.480 0.408
Sand 0.54 50
0.524 0.299 1550
1
1600
0.390 0.361
Sand 0.68 25
0.664 0.468 750
28000.618 0.482

Figure 5: Tributary area for each spring
Es

From elastic theory, the following relationships can beobtained :

MN/m2
50

∆l =

18.75

σ

C

(3)

σ = Applied stress.
C = Elastic constant of the spring

25
53.57
AV. E s=36.83C=

Therefore a constant value of the compression modulus E s= 30MN/m2 is adopted for the analysis since preliminary
structural analysis have shown that the possible extreme
values of E s have no significant influence on static forces,
only the deformation of tunnel lining will increase slightly
if E s is varied from 30MN/m2 to 15MN/m2.

σ=

Es
R

(4)

Deformation of the springPP=
A b ∗ R ∗ ∆ΦP= Load of the spring.
A= area of soil under consideration.
R= Radius of tunnel.
b=breadth of influence = 1.0 m in this case.
Substituting 4 and 5 in to 3 :

328

(5)

∆l =

P∗R
P
=
R ∗ ∆Φ ∗ E s E s ∗ ∆Φ(6)

E s = Compression modulus of soil .
According to Hook's law, the deformation of a springobtained as follows : see figure 5.

Table 3 presents the maximum values of :Axial forces N, shear forces V and bending moment M fora segment thickness t s = 0.3 m for both load cases LC 1
and LC 2 .

is

P
∆l σ
Strain ε =
=
= A
Es
l
Es
P
( l = 1.0m )∆l =Es ∗ A

Table (3) : summary of member force , shear and bending
moment ( maximum values )Load case
Axial force
Shear force
Moment Mz
F x (KN)
V y (KN)(KN.m)
LC 1-957.4*±149.0-4.67*-745.8±149.0±59.54LC 2-761.6*±195.09.20*-567.6±195.0-107.30(7)(8)

Equating 6 and 8PP=
E s ∗ ∆Φ E ∗ A∴ A = ∆Φ

(9)

These values are decisive for design of tunnel segments .

(10)

9. STRUCTURAL DESIGN OF SIGMENTS

7. TUNNEL LINING SIMULATION FOR FINITEELEMENT ANALYSIS

Details of the structural design can not be presented in thispaper with very limited space available .However ,detailed calculations for t=0.3m thickness for LC 1 showedthat the thickness is adequate and only temperature
reinforcement ( minimum reinforcement ) is required in
top and bottom faces of the segment in the both directions
to limit concrete cracking .
Therefore the final decision is to use reinforced concretesegmental lining with thickness t= 0.3m , since it isadequate , safe and economical [3].

Physical modeling of tunnel lining and excavation in soft
ground has played an important role in studies related tounderground structure, [6]. These models include:
analytical models using formulas in accordance withnational standards with properly selected design loads,
numerical models using finite element programs (2-D and3-D) with constitutive laws in accordance with national
standards to obtain; stresses, strains and bending
moments,[5].
The segmental tunnel lining can be represented by beams
supported by springs which represent soil and joints.
Seven joints (J1 to J7 ) are assumed at which the bending
moment is zero , but there will be axial force and shear atthese joints . Figure 4 shows the basic model for segmental
tunnel lining analysis.
The system consists of:
7 joints (1 to 7 ) , 92 elements (1 to 92) and 85 nodes ( 1 to85 ) , as shown on figure4.
In this paper, the thickness of segment is assumed t s= 0.30m , although in the original study another thickness
t s = 0.35m was also investigated internal diameter oftunnel d i =5.90m. thickness of tunnel lining t s = 0.30m .
mean dimeter to the center of lining d m = 6.20m .Mean
radius to the center of lining R m = 3.10m .
Member properties :Beams : breadth B=1.00m , cross sectional area = 0.30m2 ,moment of inertia I == 0.00225 m4 .

10. NODAL DISPLACEMENTS OF THE TUNNEL.
The approximate radial displacement (∆s) at a point
directly around a tunnel in an elastic soil is given by :

∆s =

(11)

Where : P Z = stress in the ground , ν = Poisson’s ratio , a =radius of tunnel, r= radius of point of interest.
The finite element analysis had shownthat thedisplacement ∆s= 23.6mm for LC 1 and lining thickness
t=0.30m . foe LC 2 the displacement is ∆s = 30mm whichis the maximum value . in comparison with the diameter of
the tunnel ( D = 6.20m) , this displacement is relatively
small and equal to only 0.48% . The neoprene gasket
provided on all side of each segment and the high tensile
strength friction grip bolts will provide construction
measure that prevents endangering the impermeability oftunnel lining due to those values of displacement .

Assumed Modulus of Elasticity for Concrete E con =37000MN/m2.
Springs of foundation support :Breadth of influence = 1.0m.
Length of spring = 1.0m .
Central angle== 11.25 .
Area of influence for each spring =PZ (1 + ν )a 2
E∗r

11. CONCLUSIONS
From this paper the following conclusions may be drawn :11.1 the alluvial soil deposits in Baghdad city consist of
top cohesive layers underlain by lower sand layers which
extend to large depths. The top cohesive layers are overconsolidated due mainly to shrinkage stresses and rise ofwater table , stiff to hard soils with relatively low
compressibility .

= 0.1964 m2

Moment of inertia of spring I s =0 .
Compression modulus E s = 30MN/m2.
Reinforced concrete design was carried out according to[1].
8. SUMMARY OF RESULTS OF ANALYSIS

329

11.2 precast segmental linings can be used successfully inalluvial soil deposit below water table .
11.3 this study presented the design geotechnicalparameters required for analysis and design of boredtunnel .
11.4 Two load cases were considered for the design oftunnel lining; LC1, water table assumed at ground surface,
LC2, water is assumed at bottom of tunnel, a surcharge of30 KN/m2 was assumed for both cases. The earth pressures(vertical and horizontal ) and water pressures acting on thetunnel were calculated according to these assumptions,which simulate the real construction conditions.
11.5 the values of compression modulus E s varies from18.75 MN/m2 for silt to 53.6 MN/m2 for dense sand, withan average design value for the mixed soil Es= 30 MN/m2.
11.6 the tunnel lining can be simulated as 2-D frame using
finite element analysis. The analysis carried out using theobtained values of axial forces, shear forces and bending
moments showed that reinforced concrete segments of
0.3m thick. will be adequate , safe and economical for thisproject with minimum temperature reinforcement .
11.7 the maximum displacement of the lining ∆s = 30mm
for LC2, which is equivalent to 0.48% of the tunnel
diameter. This displacement is considered relatively small
and will not affect the water proofing of the lining.
REFERENCES
1. British Standard Institute,BS 8110(1997) code ofpractice for design of concrete structures, BSI,
London.
2. British Standard Institute,BS 5930 (1999) code
practice for site investigation,BSI,London.
3. British Tunneling Society, ICE(2004). Tunnel
lining design guide Tomas Telford, London.
4. Chehade,F.H and Shahrour,I,(2008). Numericalanalysis of the interaction between twin tunnels,
Influence of the relative position and construction
procedure. Tunneling and underground spacetechnology vol.23,Iss.2.
5. Federal Highway Administration FHWA(2009),
Technical manual for design and construction ofroad tunnels-civil elements report No.FHW-NHI10-034,Washington D.C.20590.
6. Meguide,M.A,Saada.O,Munes,M.A&Matter,J.D(
2008),Physical Models of Tunnels in SoftGround,a Review. Tunneling and UndergroundSpace Technology vol.23,Iss.2.
7. National Center for Construction Laboratories
NCCL(1981),Site Investigation Report for
Baghdad Metro Project,Baghdad Rapid Transit
Authority,BRTA,Baghdad,Iraq.
8. Nunes,M.A.,&Meduid,M.A(2009) A Study of theEffect of Overlying Soil Strata on the Stress
Developing in Tunnel Lining . Tunneling andUnderground Space Technology vol.24,Iss.6.

330

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect Of Silt Content On The Anisotropic Behaviour Of Silt-Sand
Mixtures2

Navid Khayat 1, Abbas Ghalandarzadeh
Ph.D. Assistance Professor, Department of Civil Engineering, Islamic Azad University Ahvaz Branch, Ahvaz,
Iran. (Khayat@iauahvaz.ac.ir)
Ph.D,Associate Professor, School of Civil Engineering, University college of engineering of Tehran, Iran,(aghaland@ut.ac.ir)
P1F

P

P2F

ABSTRACTThe anisotropy of sand has not been fully studied in terms of the effects of different silt content. This study investigates
the effects on anisotropy in Tehran sand with various silt contents. The tests performed in this research include some
Hollow Cylinder Torsion Shear Tests conducted on 400kPa confining stresses. The samples had 0 ,15, 30, and 50( )percent silt contents, and the effect of the inclination angle on the major principal stress α was investigated with aparticular emphasis on silt content. The parameter "α" is considered as the key parameter that indicates the anisotropycharacteristics, and can vary from 0 to 90 . According to the results, an increase of α leads to more contractive
behaviours in the studied sands. In all samples, increasing the silt content up to 50% causes the strength to decrease.
Keywords: Sand; strength; Silt content; anisotropy;
parameter
indicatesb′ = (σ 2 − σ 3 ) / (σ 1 − σ 3 )intermediate stresses. Different principal stress directions
and b′ values can cause changes in soil behaviour. Thus,

1 INTRODUCTION
In recent years, researchers have performed extensive andcomprehensive studies on clean sand, determined different
aspects of its behaviour in terms of its structure, and havemade several general conclusions. However, the behaviourof clean sand containing silt is associated with several
complications that make it difficult to understand.
Extensive studies have focused on the effects of inherent
and induced anisotropy on the behaviour of clean sand[14]-[17]-[19] ]-[34].
Yamamuro and Lade (1998) extensively
investigated the behaviour of sands containing small
amounts of silt (7%)[30]. They concluded that when asmall amount of silt is added to the host sand, the obtained
sand-silt mixture exhibits the opposite behaviour of that ofpure sand. The sand under low confining pressure (25kPa) is completely liquefied, and the strength of sandreaches zero. As the confining stress increases, the sandwith 7% silt has more stable behaviour than that of puresand with no liquefaction under high confining pressures.
According to Yamamure and Covrt (2001), the behaviourof silt-sand mixtures with high amounts of silt (40%) are
different from those of pure sand and those of silt-sandmixtures with small amounts of fine particles [31]. In
these mixtures, like the sands with 7% silt, liquefaction
occurs completely under 25 and 50 kPa confining stresses;
however, as the confining stress increases, the behaviour
tends to be more stable. Such behaviours are similar tothat of the sand with 7% silt content, but are still different
from those of pure sand [31].
Parameter α  (the inclination of major principal stresswith respect to the depositional (vertical) direction) is used
to denote the rotation of principal stresses, andα  and b′ are selected as the key parameters for studying

the anisotropy of sands. In this regard, Hollow Cylindrical
Torsional Shear (HCTS) is the best apparatus for studying
these parameters [1]. HCTS has been used in several tests
conducted on sands by different researchers to study theeffects of parameters α  and b′ more accurately. They
generally came to the conclusion that sandy samples had
softer behaviour as the values of α  and b′ increased inundrained tests [34]. According to Symes et al. (1985),
sandy samples exhibit more contractive behaviour as theα  and b′ values increased [14]-[25].

Despite extensive studies conducted on the anisotropy
effects of clean sand, there are few studies that haveinvestigated its effects on the behaviour of sand-silt
mixtures. Among recent relevant studies, Bahadori and
Ghalandarzade (2008) studied the inherent anisotropic
effects on the undrained behaviour of sands using HCTS
[4].
In this research, the anisotropic behaviour of Tehran sandwith different silt contents in 400 kPa initial effective
confining stress are studied. A total of 12 tests are
conducted on Tehran sand with silt contents of 0%, 15%,30% and 50%. In these tests, the effects of the inclination
angle of major stress and silt content are investigated.

Material and Methods
2.1 Material properties
Tehran sand have been used in this research, which is331were set up. A computer program was utilized to control all
test procedures by means of a PID control algorithm. Data
logging device was the last part of the control system. The
sample dimensions were 12 cm in height, 6 cm in innerdiameter and10 cm in outer diameter.
The maximum inclination of principal stress ( α  ) andintermediate stress ratio ( b′ ) are considered the criticalparameters in this research. Because the torsion sensor
measures the torque (T) at any moment, the σ Zθ (shear

produced by crushing plants, This sand with greenish graycolor. The physical characteristics and grading curves ofthis sands are shown in Fig. 1 and Table 1. The plasticity
index of silt is less than 5% and can be addressed as
non-plastic soil (see Table 2).

2.2 Maximum and minimum void ratios ofsand-silt mixtures

An effective factor in the general behaviour of saturated
sands in monotonic loading is soil relative density, aparameter that shows the sample’s specifications at the endof consolidation stage. The samples prepared at low
densities showed contractive behaviour in the drained tests;
this behaviour causes excess pore water pressure inundrained tests which reduces the effective confined stress.
The change of stress-strain from the softening to hardening
state is generally related mostly to the relative density,
which is observed and reported by several researchers[5]-[7]-[8]-[16].
It is clear that adding silt to the host sand changes the
maximum and minimum void ratios. The parameters
related to soil density are needed to compare the test
results; therefore, minimum and maximum void ratios
should be obtained for different percentages of silt. In thisresearch, ASTM 4253 and ASTM 4254 are used tomeasure the minimum void ratio ( emin ) (the densest state)stress) value can be obtained using Eq (3), where ri and ro
are the inner and outer radiuses of the samples,
respectively. The value of σ Z (vertical normal stress) must
be kept constant, and therefore, the values of σ θ andσ r (the circumferential and radial normal stress,
respectively) obtained through Eqs (1) and (2),respectively, are also kept constant. Furthermore, Pi andPo (inner and outer water pressures) are computed using
Eq (5), Fν (vertical force) is found using Eq (4), andprinciple stress can be measured by equations 6 to 8 [15].
σθ = σ Z −σr = σZ −and maximum void ratio ( emax ) (the loosest state),
respectively. The obtained values are plotted in Fig. 2. For
host sand, the minimum void ratios decrease as the silt
content of the samples increases. This trend continues up to
28 percent fine content (FC); then, the void ratio increases
as the silt content increases. The minimum achieved void
ratio takes place in the fine contents, which will be called
the threshold fines content.

(1)

2σ Zθ
tan 2α1σ Zθ (cos 2α − 2b + 1)
sin 2α
3T

T

(2)


+
σ Zθ = 2  2π (ro3 − ri3 ) π (ro3 + ri3 ) (ro − ri ) σZ =FV + π ( Po ro2 − Pi ri 2 ) − Ar PoAS

σ r (ro + ri ) + σ θ (ro − ri ) Pi =
2ri


σ
(
r
r
)
+
+ σ θ (ro − ri )
r
o
i
P =
o
2ro


2.3 Testing apparatus and method

In order to study the inherent anisotropy, a hollow cylinder
torsional shear (HCTS) device was used. For better
investigation of post peak response the device can perform
both stress and strain controls.
A closed loop control was applied to the machine to handle
any type of desired stress history including major principal
stress rotation and intermediate stress ratio. FourElectrical/Pneumatic [E/P] transducers were used to exert
required pressure for the inner and outer cell pressure in
addition to the torsion and axial load pneumatic actuators.
An alternative DC motor for torsional strain control test
was prepared in order to study the post peak behaviour. In
the case of our tests, the motor was utilized since the strain
control was adopted in order to study the softening
response. The motor speed in all the tests remained
constant and equal to 4 degrees per minute. Eleventransducers were adjusted in order to measure different
parameters continuously .Three water pressure sensors,
two for inner and outer cell pressures and one for porepressure, one vertical displacement sensor to measure axial
strain, and one rotation angel sensor to measure torsionalstrain, two differential pressure sensors, two axial andtorsional stress sensors and finally two limit switch sensors

(3)
(4)
(5)

2.4 Experimental Program

Sample preparation method used in this study was dry
deposition method, while not using water sedimentation
method. That may alter the real anisotropy fabric. Dry
deposition method is able to create more uniform silty
sand samples according to Miura and Toki (1982)
suggestion which says pluviation methods create the mostsignificant anisotropy [12].The sand was deposited into
the frame through a funnel with a long tube. Carbon
dioxide ( CO2 ) and deaired water were passed through thesamples in their saturated states. The circulating time for
the CO2 considered in the tests conducted by Zlatovic
and Ishihara (1997) were 30 minutes for clean sand and 8hours for pure silt [35]. In the present study, the minimum
and maximum times were 30 and 135 minutes for clean
sand and the 50% silt mixture, respectively. After
circulation of CO2 , the deaired water entered from thebottom similar to CO2 but in the opposite direction ofgravity and was seeped into all the voids in the sample.
The saturation procedure continues as the confiningpressure increased, and the pore water pressure wasmeasured in several steps. If the B-value would exceed332

0.95, then the sample is assumed to be completely
saturated. At this time the saturation stage was completed
and the sample should be consolidated. Because all the
tests conducted in this research were of the Consolidated
Undrained (CU) type, all drainage valves must be opened
during consolidation and connected to the system. The
amount of water discharged from the sample duringconsolidation could be measured by reading the burette
numbers at the beginning and end of this stage. After
consolidation the shear stage started at which the shear
speed was four degrees per minute, the lowest speed thatcan be applied to the system. At the end of the test theporosity of the samples was measured.

3. Results

The goal of this research was to study the influence of theamount of silt upon anisotropy features in sand-silt
mixtures. For this purpose, sand were tested using
400-kPa initial confining stress and a silt content range of
0-50%.Also the Intermediate stresses parameter (b′) in all
tests was 0.5, Some information on these tests are given in
Table 3.
All test results are presented in the form of stress-strain andeffective stress path curves. According to these curves, thesample behaviours are similar to those of dense sands basedon the classification of undrained sand behaviour in mostconducted tests, according to Yoshimine et al. (1998) [34].
The significance of the strength of the deviator stressing
phase transformation point is evident in this figure.
According to Figs. 3 through 6, the sample strengths(deviator stressing phase transformation and failure points)
decrease as α  increases; this indicates that the soil
becomes softer. Increasing the silt content up to thethreshold level will cause the strengths to decrease, and thisreduction is more obvious for the 50% silt case (Figs. 6). In
these samples, the test conditions were the same, and the silt
percentage increased as the density increased, but
unexpectedly, strength did not increase. For example, in test
No.7, the value of Dr is 76, but in test No.28, whichcontains 50% silt, its value is 99, 6.

4. Discussing and studying the stress paths
4.1 Effect of inclination angle ( α  ) on the sand’s
behaviours
In Figs. 3 through 6, the strength of samples (deviator
stressing phase transformation and failure strength) is
reduced as α  increases, which indicates that the soil
becomes softer. This behaviour has been described by
Yoshimine and Ishihara (1998) for Toyoura sand, and by
Bahadori and Ghalandarzadeh (2008) for a Firoozkuh
sand-silt mixture[4]-[34].
4.2 Effect of non-plastic fines content on the sand’s
mixture behaviours
The effect of nonplastic fines on the behaviour ofsaturated sand has been studied by many researchers [6][13]-[23]-[29]-[32]. Adding different amounts of
nonplastic silt considerably changes the sand behaviour,333

and these changes are difficult to describe. Pitman et al.
(1994) concluded that when silt was added to Ottawa
sand, it became less collapsible in undrained triaxial
compression tests. However, others found that nonplastic
silt may decrease the strength (Troncoso and Verdugo,
1985; Sladen et al., 1985) [18]- [24]. Ishihara (1993) andVerdugo and Ishihara (1996) showed that the increased
silt content increased the potential for sand to exist innature in a contractive state, which includes thepossibility of flow failure or liquefaction[9]- [28].
Yamamuro and Lade (1997, 1998) and Lade andYamamuro (1997) observed that increasing the amount of
nonplastic silt in Nevada sand increased the volumetriccontractive character in both drained and undrained
triaxial tests, even when the overall density increased[30].
Among the problems involved in these studies is theimpossibility of fixing at least one of the parameters insand-silt mixtures. In the present study, the method
proposed by Bahadori and Ghalandarzadeh (2008) [4]
was applied to address this problem. Based on thismethod, if the same conditions are considered in sampling
and testing, then increasing the silt increases the densityand decreases the strength. Thus, while the sample
densities are equal, the sample with higher silt content has
lower strength.
The sample strengths decreased in the silt-sand mixtures as
the silt content increased (Fig. 7). The decreasing strength
as silt is added to the host sand can be explained by
Tavanayagan’s classification (1999) (Fig. 8) [21].
According to Fig. 9, the void ratios of host sands are tested
in the range of Tavanayagam’s classification case (i);however, with 15% silt content, the intergranular void
ratio (ec ) approached emax, HS (Fig. 9), which is moresimilar to that observed in Tavanayagam’s classificationcase (ii). In this latter case, the finer grains may support acoarser grain skeleton that is otherwise unstable. Thesegrains act as a load transfer vehicle between some of thecoarse grain particles in the soil matrix, while theremaining fine particles fill in the voids. In this case, theshear strength is derived from a combination of friction
along coarse grain contacts and fine particles. In thisstructure, ec increased, which decreased the strength of
the samples. This behaviour is evident as the silt
percentage in the mixtures increases and leads to anincrease in relative density. Increasing the amount of silt
content to 30% caused the intergrain void ratio, the relative
density does not increase because its threshold percentage
was 28%. Thus, an inter-fine-particle void ratio is used to
explain its behaviour, which is similar to case (iv-2) and isshown in Fig. 9. Moreover, the inter-fine-particle void
ratio in the mixture with 30% silt content is greater thanthat of a sample with 15% fine particle content. When thesilt content is increased to 50% in all mixtures, e f issmaller than ( emax, HF ) ,and ( FCth < FC ); thus, their
behaviours are similar to Tavanayagan’s classificationcase (iv-2). In this case, the fine particles carry the contact
and shear forces, while the coarse grains act as reinforcing
elements embedded within the finer grain matrix. In thissituation, the main bearing structure changes from sand tosilt, and because the strength of silt is less than that of sand,increasing the silt up to 50% decreases the strength of the
mixture.
5. Conclusions
1) As α  increases, the strengths of the samples of hostsand and silt-sand mixtures decrease, i.e., the soil becomes
softer. Such behaviours have also been observed by
Yoshimine et al. (1998) and Bahadori and Ghalandarzadeh(2008) [34]-[4].
2) As the silt content in mixtures increases, the relative
density also increases, but this increase does not increase
the mixture strength. An increase in relative density is oneof the most marked phenomena indicating that the index of
the relative density (or void) is not a suitable index forevaluating the behaviour of silt-sand mixtures.
3) In all mixtures, increasing the silt content up to 50%
causes the strength to decrease. This decline in strength
may be justified by parameters ec , e f , which are defined()

by Tavanayagam (1999) [21].
Notation
The following symbols are used in this paper:
α : the inclination of the major principal tress with respect
to the vertical directionb′ : Intermediate stressesb : Portion of fine grains that contribute to active
intergrain contacts
D : Diameter of coarse grains
Cc : Coefficient of curvature

Cu : Coefficient of uniformity
D50 : Medium grain sized : Diameter of fine grains

ec : intergranular void ratioe f : Inter-fine-particle void ratio
emax , emin : Maximum and minimum void ratios,

respectively

emax,HF , emin,HF : Maximum and minimum void ratios ofhost fine particles, respectively
FC : Fine particle content (%)
FCth , FC L : Threshold and limiting fine particle content,
respectively(e1 − e3 ) / 2 : deviatoric axial strainpc′ : initial effective confining stresspi , po : inner

and

outer

water

pressures,

respectively

ri , ro : inner and outer radius, respectively

334

6. REFERENCES[1] Anhminh, N. (2006). An investigation of theanisotropic stress-strain-strength characteristics ofEocene clay. Thesis for the degree of Doctor ofPhilosophy and for the Diploma of the Imperial
College.
[2] ASTM D4254. Standard Test Methods for Maximum
Index Density and Unit Weight of Soils Using aVibratory Table. ASTM Standard, (2006).
[3] ASTM D4254. Standard Test Methods for Minimum
Index Density and Unit Weight of Soils andCalculation of Relative Density. ASTM Standard(2006) e1.
[4] Bahadori, H. Ghalandarzadeh, A, & Towhata, I.
(2008). Effect of non plastic silt on the anisotropic
behaviour of sand. Soils and Foundations 48, No. 4,
531-546.
[5] Bishop, A. W., (1971). Shear strength parameters forundisturbed and remolded soil specimens. Proceeding
for the Roscoe memorial symposium, Cambridge
University, Cambridge, Mass, pp 3-58.
[6] Bouckovalas, G. D, Andrianopoulos, K. I. &Papadimitriou, A. G., (2003). A critical stateinterpretation for the cyclic liquefaction resistance ofsilty sands. Soil Dynamics and Earthquake
Engineering, No. 23, 115-125.
[7] Castro, G. (1969). Liquefaction of sands. Harvard Soil
Mechanics Series, No.81, Pierce Hall.
[8] Castro, G., Poulos, S.J., France, J.W., and Enos, J.1.
(1982). Liquefaction induced by cyclic loading.
Geotechnical engineers Inc., Report submitted to NSF.
[9] Ishihara, K.(1993). Liquefaction and flow failure
during earthquakes. Geotechnique 43, No.3, 351-415.
[10] Khayat, N. (2010). The investigation of the effect ofinherent and induced anisotropy on the undrained
behavior on Silt-Sand mixture. Thesis for the degreeof Doctor of Philosophy of the Department of Civil
Engineering, Science and Research Branch, Islamic
Azad University,Tehran, Iran.
[11] Lade, P.V. and Yamamuro, J.A. (1997). Effects of nonplastic fines on static liquefacltion of sands. Candian
Geotechnical Journal, 34.No. 6, 918-928.
[12] Miura. S. and Toki, S. A.(1982). Sample preparation
method and its effect on static and cyclic
deformation-strength properties of sand, Soil andFoundation 22, No. 1, 61-77.
[13] Naeini, S.A. & Baziar, M.H. (2004). Effect of fines
content on steady-state strength of mixed and layered
samples of a sand.. Soil Dynamics and Earthquake
Engineering 24, 181-187.
[14] Nakata, Y., Hyodo, M., Murata, H. and Yasufuku, N.
(1998). Flow deformation of sands subjected toprincipal stress rotation. Soils and Foundations 38, No.
2, 115-128.

[32] Yang, S.L., Sandven, R., & Grande, L. (2006).
Instability of sand-silt mixtures. Soil Dynamics andEarthquake Engineering, No.26, 183-190.
[33] Youshimine, M., Ishihara, K. and Vargas, W. (1998a).
Flow deformation of sands subjected to principal
stress rotation. Soils and Foundations 38, No.3,
179-188.
[34] Yoshimine, M. & Ishihara, K., (1998). Flow potentialof sand during liquefaction. Soils and Foundations
38,189-198.
[35] Zlatovic, S, & Ishihara, K., (1997).Normalized
behaviour of very loose nonplastic soils: effects offabric. Soils and Foundations 37, No.4, 47-56.

[15] Pradhan, B. S., Tatsuoka, F. and Horii, N. (1988).
Simple shear testing on sand in a torsional shear
apparatus, Soils and Foundations 28, No. 2, 95-112.
[16] Seed, H.B. and Idriss, I.M.(1971). Simplified
procedure for evaluating soil liquefaction potential.
Geotechnical Engineering Journal 97, No.9,
1249-1274.
[17] Sivathayalan, S. & Vaid, Y.P. (2002). "Influence ofgeneralized initial state and principal stress rotationon the undrained response of sands", Canadian
Geotechnical Journal;39:63-76.
[18] Sladenn J. A., D'Hollander, R. D. and Krahn, J.
(1985a).Back analysis of the Nerlerk bermliquefaction slides. Canadian Geotechnical Journal
22, No.4, 564-588.
[19] Symes, M.J., Gens, A. and Hight, D.W.(1984).
Undrained Anisotropy and Principal stress Rotation in
Saturated Sand. Geotechniqu 34, No. 1, 11-27.
[20] Symes, M.J., Gens, A. and Hight, D.W.(1985).
Discussion. Geotechnique, 35, No. 1, 78-85.
[21] Thevanayagam, S. (1999). Effect of fines and cofining
stress on undrained shear strength of silty sands.
Journal of Geotechnical and Geoenvironmental
engineering 124, No.6, 479-491.
[22] Thevanayagam, S., Shenthan, T., Mohan, S. andLiang, J. (2002). Undrained fragility of sands, silty
sands and silt. ASCE, Journal of Geotechnical andGeoenvironmental engineering 128, No.10, 849-859.
[23] Thevanayagam, S. & Martin, G.R. (2002).
Liqueffaction in silty soils-screening and remediation
issues. Soil Dynamics and Earthquake Engineering,No.22, 1035-1042.
[24] Troncoso, J.H. & Verdugo .R.(1985).Silt content anddynamic behaviour of tailing sand .Proceeding of 11th
JCSMFE,
Balkema,Rotterdam,The
Netherlands,pp.1311-1314.
[25] Uthayakumar, M. & Vaid, Y. P. (1998). Staticliquefaction of sands under multi axial loading.
Canadian Geotechnical Journal 32, No.2, 273-283.
[26] Vaid ,Y.P., Sivathayalan, S.,Uthayakumar , M. And
Eliadorani, A. (1995a). Liquefaction potential ofreconstituted Syncrude sand. 48th Canadian
Geotechnical Conference 1, pp.319-328.
[27] Vaid ,Y.P., Uthayakumar, M.Sivathayalan, S.,
Robertson, P.K And Hofman,B. (1995b). Laboratory
testing of Syncrude Sand . 48th Canadian
Geotechnical Conference 1, pp.223-232.
[28] Verdugo, R. & Ishihara, K. (1996). The steady state ofsandy soils. Soils and Foundations 36, No.2, 81-92.
[29] Xenaki, V.C. & Athanasopoulos, G.A. (2003).
Liquefaction resistance of sand-silt mixtures: anexperimental investigation of the effect of fines. Soil
Dynamics and Earthquake Engineering, No.23,
183-194.
[30] Yamamuro , J. A. & Lade, P.V. (1998). Steady-state
concepts and static liquefaction of slity sands. Journalof Geotechnical and Geoenvironmental engineering
124, No.9, 868-877.
[31] Yamamuro , J.A. & Covert, K.M. (2001). Monotonic
and Cyclic Liquefaction of Very Lose Sands withHigh Silt Content. Journal of Geotechnical andGeoenvironmental Engineering, ASCE127, No. 4,314-324.
335

Table 1: Physical characteristics of sands
Sand type

Cc

Cu

FC (% )

D50(mm )Tehran sand

0.59

2.55

0.03

0.106
Fig. 2. Minimum and maximum void ratios for Tehran
sand-silt mixtures
U

Table 2: Physical properties of silt
Plastic limit

Liquid limit

Plasticity index

(%)

(%)

(%)

22

24

U

2

Table 3. Characteristics of the tests
Test

Sand

FC

No.

type

(%)

7

T

8

α

e

Dr (%)

0

15

0.711

T

0

30

9

T

0

16

T

17

(e ) (e )
f

c

76.0

1.83

0.71

0.712

75.8

1.83

0.71

60

0.718

74.3

1.85

0.72

15

15

0.600

83.0

1.25

0.84

T

15

30

0.610

80.7

1.27

0.85

18

T

15

60

0.606

81.6

1.26

0.84

25

T

30

15

0.560

89.3

0.98

0.92

26

T

30

30

0.563

88.6

0.98

0.93

27

T

30

60

0.567

87.7

0.99

0.93

28

T

50

15

0.530

99.6

0.76

0.88

29

T

50

30

0.534

98.8

0.77

0.88

30

T

50

60

0.537

98.2

0.77

0.89

subjected to an initial effective stress of 400 kPa (FC=0%).

Fig. 4. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa

100

percent by weight finer (%)

Fig. 3. Stress path and stress-strain curves in Tehran sand90

(FC=15%).

80
70
60
50
40
30
20
10
0
0.01000.10001.000010.0000

Grain Size (mm)Fig. 1. Grain size distribution curves of Tehran sandFig. 5. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa(FC=30%).

336

Tehran Sand
2.5

2.0

Void Ratio

e-max
e-min

1.5

0%
15%
1.0

30%
50%

0.5

0.0
0

20

40

60

80

100

Silt Content (%)

Fig. 9. Intergranular matrix phase diagram for Tehran Sand

Fig. 6. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa(FC=50%).

Fig. 7. Stress path and stress-strain curves in Tehran sandswith different silt contents at α = 15.

Fig. 8.Intergranular and Interfine Matrix Phase Diagram(Thevanayagam,1999) [18]337

Optimizing the twin tunnels position for reduction the surface groundmovements
Mehdi Mokhberi1, Seyed Ali Farjam2
1. Assistant Prof., Department of civil engineering, Islamic Azda University, Estahban Branch
2. M. Sc. Student, Department of civil engineering, Islamic Azda University, Estahban Branch

Abstract:
Increasing in the city’s population and the necessity to access to various locations has led to use ofunderground structures in urban planning. In the different cities, the twin tunnels has utilized in urban
transportations. This type of tunnel design effected the urban facilities specially caused the settlements ofroads and structures. In the researchers studies, the tunnel deformations, ground moments and tunnel
stabilities were considered. For evaluation the effect of tunnel positions on surface settlements a numerical
study of twin tunnels in Shiraz City were studied. According to different type of soils and the tunnel
positions, the ground deformation was analyzed and discussed. Three different configurations are taken into
consideration, side-by-side tunnels, piggyback tunnels and angular-offset tunnels. Empirical correlations,derived from extensive field data, are used to calculate ground settlements caused by twin bored tunnel.
Non-linear finite element analysis is used for these situations. The use of superposition was tested using thenon-linear analysis to check whether or not its use with empirical methods is appropriate. The results
confirm that is a clear difference in the magnitude and distribution of ground movements in tunnel positions.
Key words: twin tunnel, underground, tunneling, settlements, ground movments1.

the soil settlement and that the design of twin tunnels
requires numerical analysis associated to monitoring
during the design and construction [4]. The constructionof the first tunnel may significantly affect the soil
conditions: reduced confinement, stress release andreduction of the strength parameters of the soils.
Consequently, the second tunnel will be excavatedthrough a different material and the induced settlements
related to the second tunnel will be generally greater [5].
many twin tunnel case histories do not usually havesufficient data points in critical positions and are usually
given as a total profile and do not show the individual
contributions made by each tunnel. Quality data that does
fit this description has been reported [6], [7]. Due to thelack of data and through advancements in computing
power, researchers have been aiding understanding of
deformation of above twin tunnels, using the numericalmethods [1], [4], [6], [8].

INTRODUCTION

Underground transportation facilities in urban areas havea significant importance on the citizen lives as they
provide fast and safe transportation services. Therefore,
the usage of metro tunnels has improved in recent years
all over the world. Constructing multiple tunnels
provides increased engineering challenges e.g.
excavation of twin metro tunnels side by side or
piggyback requires much more attention rather than aconstructing single tunnel. Ground surface settlement is
an inevitable consequence of excavating and constructing
tunnels [1].
In order to evaluate the multiple tunnel behavior, anumber of studied were carried out. The studies focusedon experimental, empirical and numerical methods. Manysemi-empirical equations have been introduced byvarious authors for predicting the movements above
single [2], [3]. The equations have been shown overmany years, when compared to case history data of single
tunnels, to make accurate predictions of the vertical andhorizontal displacements. Furthermore numericalmodeling and in situ observations were used to analyze
the interaction between twin tunnels. Results show that insome configurations, the interaction could largely affect

Unfortunately, the resulting magnitudes of settlements
produced using these numerical methods do notaccurately reproduce the actual magnitudes ofdisplacement found at full scale. Hence they cannot be
used directly to improve the empirical equations. In thisreason, this paper presents 3D numerical analysis
conducted to investigate the influence of twin tunnel
spacing on the surface settlement and internal forces338

resulting from the tunnel excavation. Analysis was
carried out for three different tunnel situation anddifferent twin tunnels spaces.

2.

profiles. Using the finite element program analyzed thesoil to evaluate the ground surface deformation. This
research has performed to evaluate the tunnels behavior
which constructed in the different class of soils.
2.1

MATERALS AND METHODS

Soils and lining properties

In order to use the Plaxis software the soils and lining
properties are defined. 5 types of Shiraz soil properties
and 35 centimeters thickness concrete fragments havebeen taken. The Mohr-Columb elasto-plastic criterion
introduces the soil behavior. The data used in the analysis
have listed in tables I and II respectively.

In this study the analysis of the three types of twin tunnel
configuration have considered; side-by-side, piggyback
and angular-offset. Fig. 1 illustrated the tunnels status.
Common practice was carried out to superimpose theindependent settlements to get the final settlement

Table I. the data collected from different location of shiraz Metro soil

Soil
Locationclassification

Unit weight (γ)( KN )

1

GM-GP

Moa’li Abad

21

50

2

GP

Qasr Dasht

19

3

SM-SC

Farhang Shahr

4

CL-ML

5

CL

SPT

W%

γ sat

E

C

φ

( KN )( KN )
m2( KN )(degree)
m2

5

22

60000

40

5

15

15

20

40000

30

10

18

40

10

19

25000

25

15

Khak Shenasi

17

12

12

19

30000

20

20

Moshir Fatemi

20

25

15

22

20000

10

30

m3

m3

Table II. Properties of concrete lining fragments
Parametrs

amountunit

EA

1.4×107

KN/m

EI

1.43×105

kNm/m2

D

0.35

m

w

8.4

KN/m/m

υ

0.15

-

P

P

P

Fig. 1. the schematic position of twin tunnels

339

2.2

Surface effect

2.2.2

In order to recognize the twin tunnel settlements, surface
deformation for each tunnel has founded separately. An
analysis has been performed for twin tunnels and theresults measured. Following discussed the obtained
results:
2.2.1

Piggyback tunnels

The Piggyback tunnels analyzed in 8, 16, 20, 24, 32, 40
and 60 meters distances. . The water table position is 2
meter above the tunnel lining. Figs 8 and 9 shows theeffect of vertical distance on the subsurface settlements.
The results indicate that the piggyback tunnels have notinterface significantly with the maximum settlements.
The surface settlement is considerably depending on
upper tunnel settlements.

Side-by-side tunnel

Figs 2 to 7 shows the effect of distance on the subsurface
settlements. The 6 meters diameter tunnels are buried in
depth 15 meters. The tunnels spaces are 8, 16, 20, 24, 32,
40 and 60 meters respectively. The water table position is
2 meter above the tunnel lining. The settlements are
depending to the tunnels spaces. The closest tunnels havemore effect and larger settlements. Keeping a way fromeach other, the settlement is decreasing. The maximum
settlement is occurred in distance 8 from the center of
each tunnel. The settlement is remains constant between
8 to 16 meters and decreases after 16 meters from centerof tunnels. In the 60 meters distance the tunnels
settlements have no superimpose effects.

2.2.3

Angular-offset tunnels

For the Angular-offset tunnels analyze the 25, 45 and 65
degree offset-angular have been considered. The tunnels
have 8 and 14 meters offset distances and the soil class is
CL. The results show a linear decay with spacing
between two tunnels. Fig. 10 confirms that thesettlements increases from 25 to 65 degree offset. The 25
degree twin tunnel has less settlements and the 65 degree
offset have maximum superimpose settlements. The
maximum amount of settlements has occurred in thecenter
point
of
twin
tunnelsFig. 2. the total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=8 m

340

Fig. 3. the total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=16 m

Fig. 4. The total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=20 m

Fig. 5. the total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=32 m

341

Fig. 6. The total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=40 m

Fig. 7. the total settlements in side-by-side tunnels, tunnel diameter=6 m; tunnels spaces=60 m.

Fig. 8. the total settlements in angular-offset tunnels, tunnel diameter=6 m; spaces=8 m

342

Fig. 9.The total settlements in angular-offset tunnels, tunnel diameter=6 m; spaces=14 m

Fig. 10. The settlements in piggyback tunnel, tunnel diameter=6 m

3.

•

CONCLUSION

The research has carried out to recognize the effect oftwin tunnel situation on ground surface settlements.
Using the finite element method analysis the following
results could be obtained:•••

In the single tunnel, Settlements decreases withincreasing the distance from the tunnel center;
while the settlement have maximum amount in 8
to 16 meter from the center of each tunnels in
the twin tunnels, due to superimpose effect.
In spite of soil saturation, the sand layers haveconsiderable settlements than the other soils
type.
Settlement is increases with increasing theoffset-angular degree, and it has the maximum
amount in 60 degree alignments.

4.

The piggyback tunnels settlements is due toupper section settlements and the lower parts
have least effect on superimpose settlements.

ACKNOWLEDGEMENTS

This paper was funded by a part of the M. Sc. Researchin the Islamic Azad University, Estahban Branch. The
authors are grateful to the head of University, research
management and the head of the civil engineering
faculty. They are thankful to Shiraz Urban Railway
Organization (SURO) for cooperating in geotechnical
data.

343

5.

of the relative position and construction procedure."
Tunnelling and underground space technology., Vol. 23,
PP. 210-214.

REFERENCES[5] Guglielmetti, V, Piergiorgio, G, Mahtab, A and Xu Sh,
2007. Mechanized tunneling in urban areas, Taylor andFrancis Group, London, UK.

[1] Addenbrooke, T. I., Potis, D. M. 2001. “Twin tunnel
interaction: Surface and subsurface effects”. Int. Jour. of
Geomechanics, 1 (2), 249-271.

[6] Cooper, M.L., Chapman, D.N., Rogers, C.D.F., 2002.
“Prediction of settlement in an Existing Tunnelcause by theSecond of Twin Tunnels. Design of structures” 2002,
Transportation research record 1814, pp.103-112.

[2] O'Reilly, M.P. and New, B.M. 1982. "Settlements
above tunnels in the United Kingdom - their magnitude
and prediction". Proceedings of Tunneling' 82
Symposium. Institution of Mining and Metallurgy.
London. pp. 173 - 181.

[7] Nyren, R.,1998. “Field measurements above twintunnels in London Clay”. PhD theses, Imperial College[3] Peck, R., 1969. “Deep excavation and Tunneling insoft ground" In proc 7th int conf soil Mechanics and Found
engineering, Mexico., pp. 225-231.

[8] Chapman, D.N., Rogers, C.D.F., Hunt, D.V.L., 2003.
“Investigating the settlement above closely spaced multiple
tunnel constructions in soft ground”. Proc of World Tunnel
Congress 2003, Amsterdam 2003, vol 2. pp.629-635.

[4] Chehade, F., and Shahrour, I., (2008). "Numericalanalysis of the interaction between twin-tunnels: influence344

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Evaluation of Pile Lateral Capacity in Clay Applying
Evolutionary Approach
Alkroosh1 I., Nikraz2 H
Curtin University, Perth, Australia
ABSTRACTThis paper presents the development of a new model to predict the lateral capacity of piles inserted into clayey soils andsubjected to lateral loads. Gene Expression Programming (GEP) has been utilized for this purpose. The data used fordevelopment of the GEP model is collected from the literature and comprise 38 data points. The data are divided into
two subsets: Training set for model calibration and independent validation set for model verification. Predictions fromthe GEP model are compared with the results of experimental data. The model has achieved a coefficient of correlation,r, of 0.95 for training and validation sets and average prediction ratio (APR) of 0.97 and 1.04 for training and validation
sets respectively. The results indicate that the GEP model performs very well and able to predict the pile lateral capacity
accurately.
Keywords: Pile; Capacity; GEP; Training and validation

1. INTRODUCTION
Geotechnical engineers often recommend piles asfoundations to support the proposed superstructure subjectedto lateral loads. Hence, pile lateral carrying capacity isrequired to be evaluated.
Several researchers have attempted evaluation of pile lateralcapacity based on analytical solutions [e.g. 1], semiempiricalsolutions [e.g. 2, 3] and finite element solutions [e.g. 4].
However, because of the nonlinearity of the soil behaviour
and the variability of soil properties, the proposed methods
have achieved limited success in terms of giving accurate
prediction of pile lateral capacity.
In this respect, artificial intelligence techniques may be more
efficient. Recently, several attempts have been made to use
artificial neural networks (ANNs) for modeling the axial
capacity of pile foundations [e.g. 5, 6, 7] and lateral capacity[e.g. 8]. The modelling advantage of ANNs over traditional
methods is the ability of ANNs to capture the nonlinear andcomplex relationship between the bearing capacity and thefactors affecting it without having to assume a priori formulaof what could be this relationship. However, the main
shortcoming of ANNs is the large complexities of the
network structure, as it represent the knowledge in terms ofweight matrices together with biases that are not accessible tothe users [9].
In this regard, the genetic programming (GP) may represent
better alternative. The main advantage of the GP over the
ANNs is the ability to provide the relationship between a setof inputs and the corresponding outputs in a simple
mathematical form accessible to the users. Recently, the GP
has been found successful in solving several problems in thefield of engineering [e.g. 10, 11].

In this paper, the lateral capacity of piles in clayey soils has
been correlated with undrained shear strength and load
eccentricity using a developed version of genetic
programming that is gene expression programming (GEP).
Recently, GEP has been applied with success in solving
engineering problems [e.g. 12, 13, 14, 15, 16a, 16b]
The objectives of this paper:
1. Applying the GEP technique for modelling thelateral load capacity of pile foundations embedded inclayey soils.
2. Evaluating the performance of the GEP model bycomparing its predictions with experimental data.
3. Measuring the accuracy of the GEP model via
statistical analysis.
2. OVERVIEW OF GENE EXPRESSION PROGRAMMING
Gene Expression Programming is an instance of an
Evolutionary Algorithm from the field of Evolutionary
Computation, invented by Ferreira [17] as a global
optimization algorithm. It has similarities to other
Evolutionary Algorithms such as the Genetic Algorithm aswell as other Evolutionary Automatic Programming
techniques such as Genetic Programming. Similar to the GAs,
the GEP utilizes evolution of computer programs (individuals
or chromosomes) that are encoded linearly in chromosomesof fixed length and likewise the GP the evolved programs are
expressed nonlinearly in a form of expression trees (ETs) ofdifferent sizes and shapes. However, the GEP implements
different evolutionary computational method.
The GEP distinct itself from GAs in that the evolved solutions
are expressed in forms of parse trees of different sizes andstructures and unlike GP genetic variations are performed on
chromosomes before they are translated into ETs. The GEP
chromosomes are composed of multiple genes, each gene is
encoded a smaller sub-program. Every gene has a constantlength and includes a head that contains functions and345

terminals, and a tail that composes of terminals only. The
genetic code represents a one-to-one relationship between thesymbols of the chromosome; the functions or terminals. The
process of information decoding from chromosomes toexpression trees is called translation which is based on sets ofrules that determine the spatial organization of the functions
and terminals in the ETs and the type of interaction (link)
between the sub-ETs [17].
The modelling process of GEP begins with random
generation of chromosomes of initial population. Eachindividual chromosome is expressed and its fitness isevaluated through the fitness function which measures how
good the individual is at competition with the rest of the
population. The best individuals are kept for modifications
which are performed by the genetic operators such as
mutation and recombination. New offspring of chromosomeswith new traits are generated and used to replace the existing
population. The individuals of the new generation are thensubjected to the same developmental process which isrepeated until stopping criteria are satisfied
3. DEVELOPMENT OF THE GEP MODEL
In this work, the GEP model is developed using thecommercial available software package GeneXproTools 4.0[18]. The data used for the model development are collectedfrom the literature and comprise 38 data points of piles
inserted in clayey soil reported by Rao and Kumar [19] andfound in Das and Basudhar [8]. The piles have different sizes
with diameters ranging from 6.35 mm to 25.4 mm and lengths
from 130 mm to 300 mm. In order to accurately predict thepile lateral capacity, the significant factors that influence thecapacity need to be identified and presented to the GEP as
input variables. These include the pile geometry, load leaver
arm and soil properties. The pile geometry is represented by
the pile diameter, D, and pile embedment length, L. The load
leaver arm is represented by the eccentricity, e. The soil
properties are represented by the undrained shear strength,
S u. . The lateral pile capacity, Q u , is the single output.
3.1 Data Division
The next step in development of the GEP model is the data
division. In this work, the data are randomly divided into two
statistically consistent sets, as recommended by Masters [20]and detailed by Shahin et al. [21]. This includes a training set
for model calibration and an independent validation set formodel verification. In total, 29 data points (75%) of theavailable 38 data points were used for training and 9 data
points (25%) for validation. The statistics of the data used forthe training and validation sets are presented in Table 1,
which includes the mean, standard deviation, maximum,
minimum and range. It should be noted that, like all empiricalmodels, GEP performs best in interpolation rather thanextrapolation, thus, the extreme values of the data used were
included in the training set.

Table 1 GEP model input and output statistics
Model
Statistical parameters
variableand data
Mean SD*
Max.* Min.* Range
sets
Pile diameter, D (mm)Training
18
7
33
6
27
Validation
17
4
25
12
13
Pile embedment length, L (mm)Training
282
50
300
130
170
Validation
269
63
300
132
168
Load eccentricity, e (mm)Training
45
14
50
0
50
Validation
44
17
50
0
50
Undrained shear strength, Su (kPa)Training
10
10
39
3
35
Validation
10
12
39
3
35
Pile lateral capacity, Qu (N)Training
77
40
225
30
196
Validation
62
29
128
35
93
* SD: Standard deviation; Max: Maximum; Min: Minimum
3.2 Modelling Attempts
The success of the modelling process using GEP techniquedepends significantly on the design of the model structure. In
this, the optimal model parameters are determined to ensurethat the best performing model is achieved. In the search for amodel using the GEP, the number of chromosomes,
chromosome structure, functional set, fitness function,linking function and rates of genetic operators play importantrole during modelling process and choosing suitable rates ofthese parameters can reduce modelling time and effort andproduce a robust solution.
In this work, the trial-and-error approach was used todetermine the values of setting parameters. This approach
involved using different settings and conducting runs in steps.
During each step, runs were carried out and the values of oneof the above mentioned parameters (with its optimal value
being searched) were varied, whereas the values of the otherparameters were set constant (i.e. number of chromosomes =30, number of genes = 3, gene’s head size = 8, functions set =+, -, ×, and /, fitness function = mean squared error (MSE),
linking function = +, mutation rate = 0.04, and generecombination rate = 0.1). The runs were stopped after fifteen
thousand generations, which were found sufficient to evaluate
the fitness of the output. At the end of each run, the MSE for
both training and validation sets were recorded in order to
identify the values that give the least MSE. The searchattempts for optimal parameters values are presented in Table
2.

346

Table 2 Input parameters used for developing GEP modelParameter
Used input
Number of chromosomes 15, 16, 17, …30
Number of genes
1, 2, 3
Head size
7, 8, 9, …12
Function set+,- , ×, ÷, √, 3 , 4 , Powervalidation sets. It can also be seen that the model has goodaverage prediction ratios, APR, of 0.97 and 1.04 for the
training and validation sets, respectively. The APR iscalculated fromFitness function
Linking function
Mutation rate
Recombination rate

where;
Q p : predicted capacity; Q m : measured capacity and n: thenumber of case records. Fig. 2 also indicates that the model
has minimum scatter around the line of equality between themeasured and predicted pile capacities for the training andvalidation sets. The results demonstrate that the developed
GEP model performs well and provides accurate predictions.

 Qp 
 / ni =1
 Qm n

APR = ∑ 

3.3 Model Formulation
As mentioned earlier, one of the advantages of the GEP is thatit presents the relationship between the input and output in aform of expression trees as shown in Fig. 1. As can be seen,
the figure illustrates the mathematical operations andinteractions between the components of the solution. This cangive insight to the nature of the relationship between the input
and the output. The trees can be easily translated and arranged
into mathematical expression as follows:L

Qp = 2D + e +
2

e +3

−
2

su
L

e2
L
−
D(D + 9.8) S u

250
Tra ining Set (r = 0.95)
Va lida tion Set (r = 0.95)200

Predicted pile capacity (N)MSE (Mean Squared Error)+, ×
0.045, 0.05, 0.055, …, 0.08
0.1, 0.2, 0.3, …, 0.7

(2)

(1)

where;
Q p : predicted pile lateral capacity; D: pile diameter; L:
embedment depth; e: eccentricity; S u : undrained shear
strength.

150

100

50

0
0

50

100

150

200

250

Measured pile capacity (N)Fig. 2 GEP model performance in training and validation sets
Table 3 Numerical evaluation of the GEP model performance
Performance
Data setmeasureTraining
Validation
Correlation
0.95
0.95
coefficient, r
Average prediction
0.97
1.04
ratio, APR
5. CONCLUSION
Fig. 1 Expression trees of the developed GEP model;
2

Sqrt = square root; 3Rt = cubic root; X = to power 2

4. RESULTS AND MODEL VALIDATION
The performance of the optimum GEP model is shown
numerically in Table 3 and is depicted graphically in Fig. 2. It
can be seen from Table 3 that the model performs well withhigh coefficients of correlation, r, of 0.95 for the training and

The results of this study indicate that the GEP model possessa good capability in predicting the lateral capacity of piles
embedded into clayey soils; the model has achieved highcoefficients of correlation, r, of 0.95 for the data used inmodel calibration and validation. The model has also low
average prediction ratio, APR, values of 0.97 and 1.04 for thedata used in model calibration and validation, respectively;
these values indicate that the model may tend to under-predictthe pile lateral capacity. The results also demonstrate that347

GEP model performs well in comparison with theexperimental data. Overall, the output of this study has
demonstrated that resulting model correlates pile lateralcapacity and undrained shear strength of soil accurately.
2 REFERENCES[1] Poulos HG, and Davis EH, Pile Foundation Analysis &Design. New York: Wiley, 1980.
[2] Hansen B, “The ultimate resistance of rigid piles against
transversal force,” Danish Geotechnical Institute,
Bulletin No 12. 1961, pp. 5-9.
[3] Brom BB, “Lateral resistance of piles in cohesive soils,”
J. Soil Mech Found Eng., ASCE, 90 (SM2), 1964, pp.
27-63.
[4] Portugal JC, and Sceo Pinto PS, “Analysis and design ofpile under lateral loads”, in Proc. 11th Int. Conf. on
geotechnical seminar on deep foundation on bored andauger piles, Belgium, 1993, pp. 309-13.
[5] Abu-Kiefa M, “General regression neural networks fordriven piles in cohesionless soils,” J. of Geotechnical &Geoenvironmental Engineering, vol. 124 (12), 1998, pp.
1177-1185.
[6] Chan WT, Chow YK, Liu LF, “Neural network: analternative to pile driving formulas,” J. Computer &Geotechnics, vol. 17, 1995, pp. 135-56.
[7] Shahin M, 2010. “Intelligent computing for modelingaxial capacity of pile Foundations,” Canadian
Geotechnical Journal, vol. 47 (2), 2010, pp. 230-243.
[8] Das SK, and Basudhar PK, “Undrained Lateral Load
Capacity of Piles in Clay Using Artificial Neural
Networks,” Computers and Geotechnics, vol. 33 (8),2006, pp. 454-459.
[9] Rezania M, and Javadi A, “A new genetic programming
model for predicting settlement of shallow foundations,”
Canadian Geotechnical J., vol. 44 (12), 2007, 1462-1473.
[10] Javadi AA, Rezania M, Nezhad MM, “Evaluation ofliquefaction induced lateral displacements using genetic
programming,” Computers and Geotechnics, vol. 33(4-5), 2006, pp. 222-233.
[11] Bayksoglu A, Gullu H, Canakci H, Ozbakir L,“Prediction of compressive and tensile strength oflimestone via genetic programming,” Expert Syst. Appl.
vol. 35 (1-2), 2008, pp. 111–123.
[12] Cevic A, and Cabalar AF, “Modelling damping ratio andshear modulus of sandmica mixtures using genetic
programming,” Expert Syst. Appl., vol. 36 (4), 2009, pp.
7749-7757.
[13] Alkroosh I, Shahin M, Nikraz H, “Modelling axial
capacity of bored piles using genetic programming
technique,” in Proc. GEO-CHIANGMIA Int. Conf.,
Thailand, 2008.
[14] Alkroosh I, Shahin M, Nikraz H, “Genetic programmingfor predicting axial capacity of driven piles,” in Proc. 1st
Int. Symp. on Computational Geomechanics. Coted’Azur, France, 2009.
[15] Alkroosh I, and Nikraz H, “Correlation of pile axial
capacity and CPT data using gene expression
programming,” Geotechnical and Geological J., 2011a.
DOI: 10:1007/s10706-011-9413-1).

[16] Alkroosh I, and Nikraz H, “Predicting axial capacity ofdriven piles in cohesive soils using intelligent
computing,” Engineering Applications of Artificial
Intelligence, vol. (25), 2011b, pp. 618-627[17] FerreiraC.
GeneExpression
Programming:
Mathematical Modeling by an Artificial Intelligence.
Portugal, Angra do Heroismo, 2002.
[18] Gepsoft, Gene Expression Programming Tool,
2002, http://www.gepsoft.com.
[19] Rao KM, Suresh Kumar V, “Measured and predicted
response of laterally loaded piles,” in Proc. 6th Int. Conf.
and exhibition on piling and deep foundation. India,
1996, 1.6.1-1.6.7[20] Master T, Practical neural network recipes in C++. In
Academic Press. San Diego, California, 1993.
[21] Shahin M, Maier H, Jaska M, 2004. “Data division fordeveloping neural networks applied to geotechnicalengineering,” J. Computing in Civil Eng., vol. 18 (2),2004, pp. 105-114.

348

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia,Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect of Reservoir Drawdown Rate on Phreatic Line Recessionin Homogeneous Earth Dams
1.
2.

R. Ziaie Moayed1, V. Rashidian2

Associate Professor, Civil Engineering Department, Imam Khomeini International University (phone: 281-837-1153; fax: 281-837-1153;
e-mail: R_ziaie@ikiu.ac.ir).
M.Sc.
Student,
Civil
Engineering
Department,
Imam
KhomeiniInternational
University(e-mail: Rashidian@ikiu.ac.ir; Vahidrashidian@gmail.com).

ABSTRACTAn embankment dam, under filled up condition develops pore water pressure within the body of the dam. If the reservoir
water is suddenly depleted, say due to the need of emptying the reservoir in expectation of an incoming flood, then the porepressure cannot get released, which causes the upstream face of the dam to slump. This condition becomes critical if thematerials of the upstream portion of the dam are not freely draining. Depending upon the value of the coefficient of
permeability of the upstream shell material and the rate of drawdown, the phreatic line in embankment can be receded
differently. In this paper a numerical analysis has been studied to evaluate the phreatic line place due to different reservoir
drawdown rate. It is observed that the phreatic line level shows a delayed response with respect to the change of reservoir
level. Also during drawdown process, the phreatic line takes the shape of concave.
1.INTRODUCTION
One of the important points in the study stages and during
construction of earth dams is seepage through the dam body.
Seepage is the continuous movement of water from the
upstream face of the dam toward its downstream face. The
upper surface of this stream of percolating water is known as
the phreatic surface. The position of the phreatic surface
influences the stability of the earth dam because of potential
piping due to excessive exit gradient and sloughing due to thesoftening and weakening of the soil mass as if it touches thedownstream slope or intersects it.Steady seepage develops
after a reservoir pool has been maintained at a particularelevation (e.g., maximum storage pool) for a sufficient lengthof time to establish a steady line of saturation through theembankment. The seepage forces which develop in the steadystate condition act in the downstream direction [1]-[3].
Another condition that may happen in earth dams is rapid
drawdown of reservoir water table. When the reservoir is
emptied, a phreatic surface or water table is formed in theembankment. The free water surface in the embankment does
not fall as rapidly as the reservoir level. A reverse hydraulicgradient is established which results in a flow in theembankment towards the emptied reservoir. The water tablein the embankment gradually recedes as water flows out
towards the toe. The hydrostatic pressure in the embankment
decreases as the water table falls, and the soil effective
stresses increase. The period immediately after drawdown of
the reservoir level is a critical one with respect to the designof the embankment. The hydrostatic pressure in theembankment causes an unstable situation as it acts towards

the free face. The effective stresses in the soil are low
therefore the full strength of the soil is not established and thefrictional resistance is low [4], [5].
The drawdown condition is a classical scenario in slope
stability, which arises when totally or partially submerged
slopes experience a reduction of the external water level. Thisis a common situation in riverbanks, subjected to changing
river levels. Flooding conditions are critical in this casebecause river levels reach peak values and the velocity ofdecreasing water level tends to reach maximum values also.
Operation of dams requires changes in water level, which
modify the safety factor against sliding of the upstream slopeof earth dams. When the reservoir level is high, hydrostaticpressures help to stabilize the slope. A reduction of waterlevel has two effects: a reduction of the stabilizing externalhydrostatic pressure and a modification of the internal porewater pressures. The second effect has traditionally received
considerable attention in dam design because it may lead to
critical conditions of the slope. The subject has been
approached from different perspectives, which have been
largely dictated by current advances in soil mechanics.
Reference [6] discusses the practical implications of rapid
drawdown and a number of case histories associated withtotal or partial failure of the upstream slope. References [7]and [8] provide further information on drawdown-induced
failures.
Current approaches to analyze drawdown are classified into
two different groups: Flow methods, which should be appliedin relatively pervious slopes and undrained methods, whichfind applications in impervious soil slopes. Methods from thefirst group concentrate on the solution of the flow problem ina situation that involves changes in boundary conditions and amodification of the initial free surface. These methods
implicitly assume that the soil skeleton is rigid and therefore
they do not consider any modification of the initial water349

pressure because of the change in total boundary stresses
imposed by the drawdown. Methods developed to handle thisproblem include flow net analysis; methods based on ad hoc
hypothesis (typically Dupuit-type of assumptions) finite
element analysis of flow in saturated soil and finite element
analysis for saturated-unsaturated flow.
The finite elements method (FEM) can be used for the stress,
seepage, and stability analysis of slopes; where nonlinear
material behavior and complex boundary, and loading
conditions can be taken into account. It has also become
possible in recent years to perform coupled analysis ofstress-induced pore pressure generation and dissipation overtime.This paper presents a numerical evaluation on phreaticline location within an earth dam in both steady-state andrapid drawdown condition. Different rapid drawdown rates
are considered and phreatic line has been evaluatedaccordingly. In this study, finite element software, ABAQUS,
is used to determine phreatic line. ABAQUS application is a
general purpose finite element program from SIMULIA Ltd.
It has a large element library and is capable of analysis ofvariety of problems. A large class of stress analysis problems
can be solved with ABAQUS.

vertical direction. Pore fluid boundary condition applied tothe model for Steady-state analysis comprise of water headover the upstream and downstream face. Potential seepage
boundary condition is assigned to drainages outlet. The
remaining boundaries are impermeable, as it is assumed that
there is no flow of water through these boundaries.
In order to simulate rapid drawdown condition, it isnecessary to perform Transient analyses. For this purpose, 15
Transient steps after Steady-state analysis have been defined.
Each step represents 1 m drawdown of reservoir water tabledaily; consequently, the pore fluid boundary condition for
upstream slope changes with steps.
In order to perform 2D finite element analyses, the model
was discretized into basic triangle elements. Figure 2 shows
the finite element mesh used in the pore fluid analyses,
consisting of 1473 6-noded isoparametric modified quadratic
triangle elements.

2. PROBLEM DEFINITION
The problem for this study is defined as in Fig. 1. The
selected example is a simplified representation of typicalhomogeneous earth dam geometry. The dam section assumed
in the present survey is a symmetric zone section withimpervious foundation. The water level height in the reservoir
and downstream are 32 m and 4 m respectively and the freeboard is 3 m; let the 1/3 be the slope of the upstream anddownstream face of the earth dam; let the top width of theembankment be 5 m. The materials properties are chosen𝑘𝑔�more close to reality. The dry density is 1800𝑚3 and the𝑚permeability of material is 0.0864 �𝑑𝑎𝑦 . The Young’s
modulus of the embankment is equal to 20 MPa. The void
ratio is considered to be 0.6.

Fig. 1. The proposed model geometry

3. ABAQUS SIMULATIONThe boundary conditions applied to the model are such thatthe whole model is restrained from moving in horizontal andFig. 2. Finite element mesh used in study
4. ANALYSIS AND DISCUSSION
In the simulation of rapid drawdown, initially asteady-state seepage analysis is performed to determine thephreatic line in steady-state condition. After that, a transient
seepage analysis is performed to obtain seepage-induced porepressures and free ground water surface for different
drawdown rates. To model variation of water level during
drawdown, linearly variable groundwater heads with time
was specified as the boundary condition in the transient
seepage analysis. The calculated groundwater flow
parameters (hydraulic heads, pore pressures, flow rates, etc.)
were later used in the deformation analysis.
Changes in pore water pressure developed immediately
after the drawdown will be exclusively owing to total stresschanges. Therefore, if an uncoupled analysis is run, the porewater pressures inside the slope will maintain their initialvalues immediately after drawdown.
Fig. 3 shows the phreatic line within the earth dam after thesteady-state condition has been reached. The dam was
supposed to be fully saturated at first. Above the phreatic line,
Sorption would occur. It means the pore pressure above thephreatic line is negative and the degree of saturation is about 5%. The seepage will reach the steady-state condition after 20
days. Fig. 4 shows the earth dam after the rapid draw down
occurred. The water level of reservoir has been drawn down
to 2 meter from 32 meter that was before reservoir rapid draw
down.

350

Fig. 3. Phreatic line before rapid draw down

Fig 7. Receded phreatic line with rapid draw down rate of
1 meter per day

Figure 4. Water level after rapid draw down
Fig. 5 to 10 depict the phreatic line recession after 30 meter
rapid draw down with rate of 0.25, 0.5, 1, 2, 4 and 8 meter perday. It can be seen from these figures that faster the rapid
draw down rate, slower the phreatic line recession. Of course
the receded phreatic line is very close to each other in all therates. It means that the rapid draw down rate has a little andnegligible effect on phreatic line recession and consequently
on upstream slope stability.

Fig 8. Receded phreatic line with rapid draw down rate of
2 meter per day

Fig 9. Receded phreatic line with rapid draw down rate of
4 meter per day

Fig 5. Receded phreatic line with rapid draw down rate of
0.25 meter per day

Fig 10. Receded phreatic line with rapid draw down rate of
8 meter per day
Fig 6. Receded phreatic line with rapid draw down rate of
0.5 meter per day

Maximum pore pressure within the dam is presented in table
1. it can be seen that the difference between the pore pressureswith different rapid draw down rates is very small and almost
omissible.
351

Table 1. Maximum pore pressure within the dam after rapid
draw down

Rapid Draw Down
Rate (m/day)
0.25

Maximum Pore Pressure(kPa)+232.50.5+244.61+251.72+255.84+258

8+259.2[2] Lambe, T. W., and Whitman, R. V. (2000). Soilmechanics, Wiley, Singapore.
[3] Neuman, S. P., and Witherspoon, P. A. (1970). “Finite
element method of analysing steady seepage with a freesurface.” Water Resour. Res., 6(3), 889–897.
[4] Sherard, J. L., Woodward, R. J., Gizinski, S. F., andClvenger, W. A. (1967). Earth and earth rock dams,
Wiley, New York.Kimura S, “Journal paper title,” J. of
Computer Science, vol. 1, Aug. 1987, pp. 23-49.
[5] Taylor, R. L., and Brown, C. B. (1967). “Darcy flow
solutions with a free surface.” J. Hydraul. Div., Am. Soc.
Civ. Eng., 93(2), 25–34.
[6] Sherard, J. L., R. J. Woodward, S. F. Gizienski, and W.
A. Clevenger (1963), Earth and Earth-Rock Dams, John
Wiley, Hoboken, N. J.
[7] International Committee on Large Dams (1980),
Deterioration of Dams and Reservoirs. Examples andTheir Analysis, Balkema, Rotterdam,
Netherlands.

5. CONCLUSION
Modern numerical methods and constitutive models enable
analysis of transient seepage in order to more realistic andeconomical design. It is shown that the simultaneous
generation of seepage and pore pressures and their dissipation
with time can be computed and the distribution of excess porepressures within the slope at any stage together the phreatic
surface can be determined.

[8] Lawrence Von Thun, J. (1985), San Luis Dam upstream
slide, Int. Conf. on Soil Mech. Found. Eng., 11, 2593–
2598.

The stability of a submerged slope during drawdown isgreatly influenced by the how fast its pore water drains. Even
though the common design practice of assuming either freely
draining (fully slow) or undrained (rapid) drawdown provides
solutions for two extreme limiting conditions, it does notreflect the field behavior realistically in majority of cases
where the hydraulic conductivity of soil and the rate and levelof drawdown combine in a manner that results in solutionsquite different than these limiting conditions.
In this paper the effect of rapid draw down rate on phreaticline recession was investigated using finite element software,
Abaqus.
It was seen that the change in rapid draw down rate has alittle effect on phreatic line recession and all the recededphreatic lines are almost similar to each other. Also there is alittle difference between the maximum pore pressure within
the dam after rapid draw down.
3. REFERENCE[1] Casagrande, A. (1937). “Seepage through dams.” J. New
England Water Works Association, reprinted inContribution to soil mechanics 1925- 1940, Boston
Society of Civil Engineers, Boston, 1940.
352

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The Effect of lime-Microsilica Stabilization on California Bearing Ratio of Silty
Soils
R. Ziaie Moayed1, Y. Daghigh2, B. Pourhadi3 and N. Sahebzamani4
1,3
Imam Khomeini International University, Iran
2
Islamic Azad University , Karaj branch
4
Islamic Republic of Iran Railways
P

P

5T

0

5TP

P

T5

0T5

5T

ABSTRACT
Silty soils have some undesirable characteristics that challenge the geotechnical engineers in practical projects. Thesesoils due to their low cohesion and friction angel cause the pavement, railways, roadways, foundation and channels to
crack and break up. In this research, the effects of lime and microsilica stabilization in a silty soil are investigated. In thisresearch, the effects of lime and microsilica stabilization in a silty soil from Ahvaz-Khorramshahr railway subgrade areinvestigated. Microsilica is one of the harmful byproducts that obtained from the factories’s chimneys that produce
silicon and ferrosilicon. Therefore to determine the best mixture of lime and microsilica to stabilize the soil, different
percentage of these materials were mixed with soil. After 7day curing time, CBR tests were conducted on specimens.
Results show that the California Bearing Ratios are increased as the content of mixed stabilizer increase. So it can beconcluded that the microsilica waste material can be successfully used to enhance silty soil resistance.
Keywords: Soil stabilization, Silty soil, Lime, Microsilica, CBR

1. INTRODUCTION
Silt is a kind of sedimentary geomaterial consisting primarilyof very fine particles, including fine sand particles, silt
particles, and some clay particles which are often less than
10% by weight. Silt is a type of transitional soil between sandand clay. A soil is defined as silt if its plasticity index is no
greater than 10 and the amount of particles greater than 0.075
mm is no greater than 50% of the total [1].
silty soils don’t consider as suitable materials in civilengineering project due to their low cohesion and friction
angel. Using the soils as a road or railway subgrade is
generally not possible without stabilization as theircharacteristics fall below the minimum required.
Consequently, stabilization is needed for this kind of soil.
Application of stabilizing agents on soils has a long history.
Cement was first used as stabilizing agent at the beginning of
the twentieth century to mix with soils and form roadmaterials in the United States. Since then, many other kindsof materials, such as lime [2] and special additives such as
Pozzolanic materials like Fly Ash [3], Microsilica [4], andRice Husk Ash [5], which are as waste material, may be used
for soil improvement. Most of the existing stabilizers likelime and cement aren’t much useful for silts, so the stabilized
silts with such kind of stabilizing agents usually cannot
satisfy the requirements of road construction. The
encountered problems mainly are lower early strength,
greater shrinkage, easy cracking, and bad water stability [6],[7].
Indeed, a successful stabilization method depends on manyfactors such as:(1) Soil type and properties; (2) stabilizing agent; (3)Stabilizer content; (4) Potential use of the stabilized soil; (5)Field mixing method; and (6) Economical considerations [8].
Therefore, new methods are still being researched to increase
the strength properties of silty soils. In this study we evaluate
the feasibility of using stabilized silt with microsilica andlime for Karaj railway subgrade in Iran.
Microsilica (or silica fume) is one of the by- product materials
which is obtained from silicon material or silicon alloy metal
factories. It was discharged into the atmosphere by thefactories smoke before the mid-1970s. Nowadays each year
nearly 100,000 tons of microsilica is produced on purpose
word wide [9]. Iran also has a large amount of microsilica
production. Although the microsilica is a waste material ofindustrial applications, it has become the most valuable
by-product among the pozzolanic materials due to its very
active, high pozzolanic property and very fine particles.
These particles are approximately 100 times smaller than the
average cement particle [9].
In previous studies, there have been many researchersinvestigating the effects of microsilica on the strength andswelling characteristics of clayey soils were investigated. It
was seen that microsilica improved the properties of clayeysoils [10]-[13]. Likewise, recently, the effects of microsilica
and lime have been investigated on CBR values of sand [9],[14]. So their effects on cohesionless soils especially siltsaren’t investigated enough yet. Therefore our aim in thisstudy is to evaluate the feasibility of using stabilized silt withmicrosilica and lime for a railway subgrade.

353

2. MATERIALS
2.1 Soil
The silts involved in this research were obtained from an areain Karaj railway project in Iran. Atterberg limits tests werecarried out according to ASTM D 4318. The soil Plasticity
Index (PI) was 2. The soil was classified as a low plasticity
soil according to the unified soil classification system ASTM
D 422 - 87. The soil name is ML according to USCS (silty soil
with low plasticity). The soil classification is shown in Fig 1.

2.3 Microsilica
Microsilica has been obtained from Ferroalloy Industrial Co(I.F.I) in Azna. The composition of microsilica mineral isshown in Table 2.
Table 2. Chemical properties of microsilica
Chimical names

Percentage

MgO
CaO
Fe 2 O 3
Al 2 O 3
SiO 2
CNa 2 O 3
SiO 2
MOP 2 O5
MoisturePH

0.5~2
0.5~1.5
0.3~1.3
0.6~1.290~95
0.2~0.4
0.3~0.5
0.04~0.08
0.02~0.07
0.04
0.01~0.4
6.6~8.8

3. EXPERIMENTAL PROGRAM

Fig. 1 Grain size distribution curve of the silty soil
2.2 Lime
Quick lime which was used in this experiment was obtained
from the industrial group Qom-Iran limestone and its
chemical composition is shown in Table 1.
Table 1- Chemical properties of lime
Chimical names

Percentage

K2 O
SO 3
MgO
CaO
Fe 2 O 3
Al2O 3
SiO 2

4
0.8
2.65
51.64
0.13
0.24
1.36

3.1 Tests Procedure
To evaluate the effects of lime- microsilica on CBR values ofstabilized silty soils, first we found soil optimum moisturefrom compaction test. Then the soil was mixed with various
contents of lime and microsilica at the soil optimum moisture.
Therefore the oven- dried soil was sieved from sieve #4 andlime and microsilica were added into them in 1, 3 and 5% forlime and 2, 5, 8 and 12 percent of dried soil weight for
microsilica. Required amount of water was added to the
mixture to obtain soil optimum moisture, beyond. The CBRtests were carried out on samples which were cured for 7 days
after 96 hours immersing according to ASTM D 1883 - 99.
3.2 Compaction Tests
To determine the soil optimum water content and the soil
maximum dry unit weight, the modified compaction tests
were carried out according to ASTM D 1557 – 91. For thispurpose, the oven- dried soil passing sieve #4 was compacted
in five layers by 56 blows with 4.5 Kg hammer from 45 cm
height in 6 inch mold according to procedure C fromrespective standard test method.
3.3 California Bearing Ratio (CBR) Tests
The California Bearing Ratio (CBR) test is one of the mostwidespread tests to determine strength and bearing capacityof base, sub- base and subgrades for use in road, railway andairfields pavements. To demonstrate the influence of lime-354

microsilica additive on the bearing ratio of the silty soil, aseries of bearing ratio tests were carried out on stabilized andunstabilized specimens. The tests were conducted according
to ASTM D 1883 – 99. The soil with different mixtures oflime and microsilica were compacted in 6" modified proctor
mold in five layers by 56 blows in per layer at the soil
optimum moisture obtained from compaction tests. For
curing the samples, they were placed in constant moisture andtemperature for 7 days. To conduct the tests in soaked
condition, they were immersed in water for 96 hours underthe 4.5 Kg (10 pound) overload according to standard test
method. The CBR tests were carried out after 20 minutes todrain the samples. Meanwhile variation of the samples
maximum unit weight and water content were determined
after the samples soaking and swelling potential changes
were measured during the soaking time.
4. RESULTS AND DISCUSSION
4.1 Compaction Tests
Compaction tests were carried out on the silty soil and the soil
optimum moisture and the soil maximum unit weight were
found to be 14.2% and 17.2 KN/m3 respectively. Compaction
tests results are drawn in Fig 2.

Fig. 3 The effect of various amount of lime- microsilica
additive on CBR values of stabilized soil
As the Fig. 3 presents, in low amounts of lime (1 and 3
percent of dry soil) increase in microsilica amount up to 8%
causes increase in CBR values and then decrease but for 5%of lime increase in microsilica amount causes increase in
CBR values. The maximum CBR value of the samples whichcontain 5% lime and 12% microsilica increases from 5.4% to393.2%. So it is seen that up to 388% increase in CBR valueof stabilized soil in compare of unstabilized silty soil.
The samples maximum dry unit weight and moisture aredetermined after samples soaking. Figs 4 and 5 show theirresults respectively. It is observed that the maximum dry unit
weight increase like CBR increase procedure and the samples
moisture are increase exactly the opposite of the samples dry
unit weight procedure. The reason for increase in the
maximum dry unit weight of stabilized samples is assessed byvoid ratio decreasing due to the additive mixture with veryfine particle size distribution.

Fig. 2 Compaction test curve

4.2 Effect of Additives on the CBR
To compare the soil resistant with different amount ofadditive, a series of samples were prepared in modified
proctor mold. The CBR tests were conducted in bothstabilized and unstabilized silty soils at the soil optimum
moisture with different amount of lime and microsilica. The
CBR value of the unstabilized soil wes 5.4%. The effect ofvarious amount of additive on CBR values of samples areshown in Fig 3.

355

Fig. 4 The effect of lime- microsilica additive on the
maximum dry unit weightFig. 5 The effect of lime- microsilica additive on samples
moisture after soaking
4.3 Effect of Additives on Swelling
The samples swelling were measured during the 96-hour of
CBR samples soaking. There were seen swelling potential
rate were decreased reverse of strength. Unstabilized soil
swelling was 0.55mm and stabilized swelling samples weredecreased up to 0.01mm.
5 CONCLUSIONSIn this research the effects of lime- microsilica additive on
silts and its utilize on railways subgrade were investigatedand following conclusions were drawn:
The experimental results showed that in 1 and 3% of lime it
was seen an increase on CBR values of samples withmicrosilica increase until 8% of microsilica and then it
increased. But in 5% use of lime the CBR values wereincreased with microsilica increase. So it sounds the amountof sufficient amount of microsilica for 1 and 3% of lime toachieve maximum pozzolanic reaction is 8% and for 5% oflime is 12%.
The results of the samples maximum dry unit weight change
is just like CBR values change procedure. The samples
maximum dry unit weight increase with CBR increase.
The moisture measured after soaking stabilized andunstabilized samples decrease exactly opposite of the samples
maximum dry unit weight change procedure.
There were also seen decreases on swelling rates of stabilized
soil by increase the additive.
Therefore, in order to related conclusions which present
considerable strength of stabilized silty soil with limemicrosilica additive in comparison of unstabilized soil,
application of lime- microsilica additive is recommended for
subgrad and even sub base of civil projects.

Quarantine of the People's Republic of China). Code fordesign of building foundation. China Architecture andBuilding Press, Beijing. (in Chinese), 2002, 262pp.
[2] Bell F.G, "Lime stabilization of clay minerals and soils."
Engineering Geology 42, 1996, pp. 223–237.
[3] Dermatas D, Meng X.G "Utilization of fly ash forstabilization/ solidification of heavy metal contaminatedsoils." Engineering Geology, 2003, pp. 377–394.
[4] Abd El Aziz, "The Effect of Silica Fume Substitution on
the Characteristics of Ordinery Portland cements Pastes
and Mortars."Civil Engineering Magazine, Vol. 24 No.
2, 2003, p. 715-725.
[5] Choobbasti A.J, Ghodrat H, Vahdatirad M.J, Firouzian S,
Barari A, Torabi M, Bagherian A, "Influence of using
rice husk ash in soil stabilization method with lime."
Earth Sci. China 2010, pp. 471–480[6] Bell F.G., "Cement stabilization and clay soils, with
examples."Environmental and Engineering Geoscience,
1995, pp.139–151.
[7] Sheng A.Q, Ma M, "Experimental study on stabilizationof subbase of bearing sand silt with low liquid limit (inChinese)". East China Highway 5, 2001, pp.42–46.
[8] Mohamedzein, Y.E., Al-Rawas, A.A. and Al-Aghbari,
M.Y. "Assessment of sand– clay mixtures for use inlandfill liners." Proceedings of the International
Conference on Geo-environmental Engineering,Singapore, December 2003, pp. 211–218.
[9] Karimi M, Ghorbani A, Daghigh Y, Kia Alhosseini S,
Rabbani P, "Stabilization of silty sand soils with lime andmicrosilica admixture in presence of sulfates" Pan- Am
CGS Geotechnical conference, 2011[10] Kalkan E, "Influence of silica fume on the desiccation
cracks of compacted clayey soils." Applied Clay
Science, 2009, pp. 296–302.
[11] Kalkan E, “Impact of wetting–drying cycles on swelling
behavior of clayey soils modified by silica fume."
Applied Clay Science, 2011, pp. 345–352.
[12] Abd El-Aziz M.A, Abo-Hashema M.A, El-Shourbagy
M, “The Effect of Lime-Silica Fume Stabilizer on
Engineering Properties of Clayey Subgrade."Engineering Conference, Faculty of Engineering,
Mansoura University, April 2004, Paper No. 96[13] McKennon, J.T, Hains N.L, Hoffman, D.C, "Method forstabilizing clay bearing soils by addition of silica andlime." Patent Cooperation Treaty (PCT), International
Application Published Under the Patent Cooperation
Treaty (PCT), Patent Classification: C09K 17/00,
Publication Number: WO 94/06884, 1994.
[14] Kalkan E, "Effects of silica fume on the geotechnicalproperties of fine-grained soils exposed to freeze andthaw." Cold Region Science and Technology, 2009, pp.
130–135.

6 REFERENCES[1] MCPRC (Ministry of Construction of the People's
Republicof
China),GAQSIQPRC(General
Administration of Quality Supervision, Inspection and356

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect of Elastic Modulus Varieties in Depth on Subgrade Reaction Modulus of
Granular Soils
Reza Ziaie Moayed1, Mahdi Ali Bolandi2
Civil Engineering Department, Imam Khomeini International University, Qazvin, Iran
ABSTRACTThe subgrade reaction modulus that indicates the relation between soil pressure and deflection is one of the most efficientparameter that is used for structural analysis of foundation members. The popular practical method for estimating the modulus
of subgrade reaction is the plate load test (PLT), that determines the ratio of load to displacement of circular plate with
15cm-75cm diameter. Afterward, the results modified with Terzaghi’s equation (1955) and became applicable for foundationengineering problems .The subgrade reaction modulus (K s ) is dependent to some parameters like soil type, size and shape offoundation, depth and stress level. One of the principle methods for determining K s is back numerical analysis. In this way, byusing Mohr-coulomb behavior model, elastic modulus (E s ) has a significant influence on obtained results. In granular soils,
due to the increase of confining stress in depth, elastic modulus is also increased. In this paper the effect of soil stress-strain
modulus (E s ) varieties in depth on subgrade reaction modulus (K s ) was investigated by using finite element analysis. The
results illustrate that in granular soils, by considering the elastic modulus varieties in depth, the subgrade reaction modulus
would become significantly greater as compared to a constant E assumed for all depth. Therefore, it may be proposed that in
subgrade reaction modulus calculation the E varieties in depth should be considered by back analysis method.
Keywords: Subgrade Reaction Modulus, Elastic Modulus, Granular Soils, Plate Load Test, Finite Element Analysis
INTRODUCTION
Soil medium has very complex and erratic mechanicalbehavior because of the Nonlinear, stress dependant,
anisotropic and heterogeneous nature of it. Hence, instead ofmodeling the subsoil in its three dimensional nature, subgrade
is replaced by a much simpler system called a subgrade modelthat dates back to the nineteenth century. The search in thiscontext leads to two basic approaches which are Winkler
approach and the elastic continuum model, which are ofwidespread use, both in theory and engineering practice.
Winkler assumed the soil medium as a system of springs
which are identical but mutually independent, closely spaced,discrete and linearly elastic and ratio of contact pressure, p, at
any given point to settlement, y, produced by load application
at that point, is given by the coefficient of subgrade reaction,
K s [1]:
Ks = p / y

There are some current methods for subgrade reaction
modulus determination like NAVFAC DM-7.1, that indicate
after plotting stress- settlement curve, q ult fined, and ratio of
0.5q ult to corresponding settlement, considered as K s [2].
According to Bowles initial tangent or initial secant line slope
can be used for estimating the modulus of subgrade reaction.
However, the slope of tangent line in applied stress may be
used to determining the subgrade reaction modulus [3].

(1)

A direct method to estimate both E s and K s is plate loading
test (PLT) that requires circular or equivalent rectangularplates. PLT provides a direct measurement of thecompressibility and bearing capacity of soil and essentiallyconsists in loading a rigid plate and determining thesettlements corresponding to each load increment. The resultsof a PLT are presented as applied contact pressure versus
settlement curves (Fig. 1).

_____________________________________________________________
1.Associate Professor, Civil Engineering Department, Imam Khomeini
International University, Qazvin, Iran, R_ziaie@ikiu.ac.ir
2.M.Sc. Student, Civil Engineering Department, Imam Khomeini
International University, Qazvin, Iran, Eng.alibolandi@gmail.com.

Fig 1. Typical presentation of results from a PLT

Many researches including Biot, Terzaghi, and Vesic haveinvestigated the effective factors and determination
approaches of Ks [4]-[6]. Geometry and dimensions of thefoundation and soil layering are assigned to be the mostimportant effective parameters on K s .
Terzaghi made some recommendations where he suggested
values of K s for 1*1 ft rigid slab placed on a soil medium;
however, the implementation or procedure to compute a357

value of K s for use in a larger slab was not specific. Biotsolved the problem for an infinite beam with a concentrated
load resting on a 3D elastic soil continuum [7]. He found acorrelation of the continuum elastic theory and Winkler
model where the maximum moments in the beam areequated. Vesic tried to develop a value for K s , except, insteadof matching bending moments. He matched the maximum
displacement of the beam in both models. He obtained theequation for K s for use in the Winkler model.
Several studies by Filonenko-Borodich, Heteneyi, Pasternak ,and others, have attempted to make the Winkler model more
realistic by assuming some form of interaction among thespring elements that represent the soil continuum [8]-[10].
One of the early contributions was that of Terzaghi, who
proposed that K s for full-sized footings could be obtained
from plate-load tests using the following equations:
For clayey soil:
K s = K p (B p / B)

different layers that E linearly change with depth. For theselayers, Mohr-Coulomb model is employed.

(4)

For sandy soil:
K s = K p [(B+B p ) / 2B)]2

(5)

Where
K s = desired value of modulus of subgrade reaction for the
full-size (or prototype) foundation
K p = value obtained from a plate-load test using a 0.3m ×0.3m(1ft ×1 ft) or other size load plateB p = plate diameter
B= dimension of foundation.

Fig 2. Elastic modulus versus depth

2.1.2 Material Properties
In present study, Sacramento River Sand properties that was
In the present paper, the effect of elastic modulus variation on given by Lade [12], dedicated to soil. Aforesaid soil
K s determination is investigated. For this purpose, the parameters are presented in Table 1.
numerical analysis are performed on four models with constant
Table 1. Sacramento River sand properties.
and various E in depth (two model for either PLT and mat
foundation) and the subgrade reaction modulus are determined
k ur
n
υ
φ
C
ѱ
ρ
using back analysis method for each model.
(o) (kPa) (o) (kg/m3)
2 ANALYSIS PROCEDURE
2.1 Material
2.1.1 Material Model
In current paper the value of E is expressed as a function ofconfining pressure σ 3 using Janbu’s [11] relationship as
E= k ur Pa (P/P a )n(6)

E = tangent elastic modulus at the mean stress P
k ur = modulus number
P a = atmospheric pressure
P= mean stress= (σ 1 +σ 2 +σ 3 )/3n= exponent modulus.
Then, by replacing P= γz (1+2 k 0 )/3, E versus depth (z)derived (Fig 2). Afterwards, the obtained curve
approximately divided to straight lines and assumed that Echanges in each part, linearly. Each part considered as

960

0.57

0.3

35

0

5

1417

2.2 Geometry
46 finite element analysis performed using Plaxis 8.5
software with axis-symmetric model. The 15-node triangularelements are used to model the soil layers and other volume
clusters. Near the edges of a loaded area where stressconcentrations are expected, mesh is refined by reducing thesize of the soil surface below the position of the loading area.
Analysis is performed under load control by a vertical
distributed load boundary condition applied to the soil surface
below the position of the loading plate. It is assumed that bothhorizontal and vertical displacements are zero for all nodes
along the bottom boundary of the mesh. On the vertical sideboundaries, the horizontal displacements have been assumed
to be zero too.
The size of axis-symmetric model in X and Y direction isconsidered greater than 2.5D and 6D, respectively, thatsatisfy Bowles recommendation [3]. Four models provided to358

analyze, two models for 0.75m diameter plate loading, andothers assigned to circular foundation with 10 m diameter. In
each of mentioned groups, one of them has layers that Elastic
modulus linearly changes in depth and other has a constant E
overall the model (Fig 3-4).

ab
Fig 5. Vertical displacement (PLT).
a: Various E, b: Constant Eab
Fig 3. Geometry of model (PLT).
a: Various E, b: Constant E

3 RESULTS
All models were analyzed (Fig 4-5) and pressure-settlement
curves for each model were plotted (Fig 6-7). Therefore, thesubgrade reaction modulus by estimating the slope of initialtangent line, were specified (Table 2).

ab
Fig 4. Geometry of model ( Mat Foundation ).
a: Various E, b: Constant E

Fig 6. Stress – Settlement curves of plate loading withvarious and constant E.

ab
Fig 5. Vertical displacement (PLT).
a: Various E, b: Constant E
Fig 7. Stress – Settlement curves of foundation loadingwith various and constant E.
359

constants”, Gosudarstvennoe izdatelstro liberaturi po
stroitelsvui arkhitekture. Moscow (in Russian), 1954.
[11] Janbu, N, “Soil compressibility as determined byoedometer and triaxial tests”, European Conference on
Soil Mechanics and Foundations Engineering,
Wiesbaden, Germany, 1963, Vol. 1[12] Lade P.V, “Elasto-plastic stress-strain theory forcohesionless soil with curved yield surfaces”, Ind. J.
Solids Structures, 1977, Vol. 13, pp. 1019-1035

Table 2. Subgrade reaction modulus for models.
Type
Plate
Foundation

K s (MN/m3)
Various E
Constant E
33.1
20.3
9.4
3.0

3 CONCLUSION
1. Considering the E varieties in depth, in granular soils,
confining pressure effect, caused the settlements
decrease. Therefore the obtained Ks become greater
as compared to constant E condition.
2. Due to significant variation of elastic modulus in low
depth, the plate load test must performe exactly at
the foundation level.
3. In order to calculate the subgrade reaction modulus
using, a specific value of E could be misleading.
4. As foundation dimensions increase, the stressinfluence depth increases subsequently. Hence theeffect of E variation in depth on Ks becomesconsiderable.
5. Based on obtained results, it may be propose that in
subgrade reaction modulus calculation the Evarieties in depth should be considered by backanalysis method.
4 REFERENCES[1] Winkler E, “Die Lehre von Elastizitat and Festigkeit (onelasticity and fixity)” . Praguc. 1987
[2] US Naval Facilities Engineering Command,1982.
[3] Bowels, J. E, Foundation Analysis and Design. fifth
edition, The Mc Graw-Hill, 1998.
[4] Biot, M. A,” Bending of Infinite Beams on an Elastic
Foundation,” Journal of Applied Science .Trans. Am.
Soc. Mech. Eng., 1937, 59: A1-7.
[5] Terzaghi, K,”Evaluation of coefficient of subgrade
reaction”, Geotechnique, 1955, Vol. 5, No. 4, pp
297-326.
[6] Vesic, A. S, ‘ Beams on elastic subgrade and the
Winkler's hypothesis”, 5th ICSMFE, 1961, Vol. 1, pp.
845-850.
[7] Biot, M. A, “Bending of Infinite Beams on an Elastic
Foundation”, Journal of Applied Science .Trans. Am.
Soc. Mech. Eng, 1937, 59: A1-7.
[8] Filonenko. M. & Borodich, M, “Some approximate
theories of the elastic foundation”, Uchenyie Zapiski
Moskovskogo
Gosudarstvennoho
Universiteta
Mekhanica, 1940, 46, pp 3-18 (in Russian).
[9] Hetenyi, M, “Beams on elastic foundations”, The
university of Michigan Press, Ann Arbor, Michigan,
1946.
[10] Pasternak, P. L, “On a new method of analysis of anelastic foundation by means of two foundation

360

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Flexural Buckling Behavior of Steel Pile
with Vertical Load in Liquefied Soil
Yoshihiro Kimura1, and Koichi Onohara 2
1 Tohoku University, 2 Nagasaki University, JAPANABSTRACT
When slender steel piles beneath buildings experience high axial compression forces as a result of vertical loads
increased by P-∆ effects with inertia forces acting on the buildings and then the soil liquefies, buckling of the piles may
occur even though they are restrained laterally by the liquefied soil.
This paper describes the evaluation of the buckling length of slender piles in liquefied soil with the unified andtriangular distributions of the coefficient of the subgrade reaction. The relationship between the buckling length of theslender piles and the soil-pile interaction is presented. Also, the elasto-plastic analyses are also conducted andinteraction between the non-linear behavior of the steel piles and inelastic soil reaction are clarified.
Keywords: Steel Pile, Flexural Buckling Length, Coefficient of Subgrade Reaction, liquefaction, Energy Method, FEM
Analysis
1. INTRODUCTION

Lateral Flexural DeformationPof Structure
When slender steel piles beneath buildings
lateral Springexperience high axial compression forces as a result
Inertia Forceof vertical loads increased by P-∆ effects with
Dead Loadinertia forces acting on the buildings and then theTensile Force due toCompressive Force due tosoil liquefies, buckling of the piles may occur even
P-D Effect
P-D Effectthough they are restrained laterally by the liquefied
Liquefied Soilsoil. For example, in the case of tall structure asshown in Fig.1, the left steel pile carries out the
Flexural Bucklingof Steel Piletensile force due to P-∆ effect, and the right one
carries out the compressive force due to P-∆ effect
Soil Reactionin addition to dead load of the structure. The
Non Liquefied SoilP
flexural buckling will occur in the right one, and
Fig.1
Conceptfor
Flexural
Buckling
Fig.2
Analysis
Model ofthen its bucking length is corresponding to the
of
Steel
Pile
in
Liquefied
Soil
Steel
Pile
in
Liquefied
Soil
thickness of the liquefied soil.
This paper describes the evaluation of thebuckling load of slender piles in liquefied soil with the
2. ELASTIC BUCKLING LOAD FOR STEEL PILE IN
unified and triangular distributions of the coefficient of the
LIQUEFIED SOIL
subgrade reaction. The equations of elastic buckling load are
2.1 Assumption of Elastic Buckling for Steel Piles
developed with the energy methods, and the relationshipIn this paper, some assumptions are applied to develop the
between the buckling load of the slender piles and theelastic
buckling load for steel pile in liquefied soil.
soil-pile interaction is presented. The elastic eigen analyses(1) The external force for a pile is only vertical load such as
are conducted to compare the solution from the equations of
Fig.2.
buckling load developed by the energy method.
(2) The boundary condition is simple support, and then the
The elasto-plastic analyses are also conducted anddisplacement of the pile at the top does not occur.
interaction between the non-linear behavior of the steel piles(3) The subgrade reaction is replaced to the elastic
and inelastic soil reaction are clarified.
horizontal springs such as Fig.2, because there is little
Finally the estimation for buckling stress is suggested
friction over depth between the soil and the pile. The
with general slenderness ratio λ c to apply to the present AIJ
coefficient of the subgrade reaction, K c is expressed
code design for the compressive steel members, and thein the following in reference to Recommendation foranalyses results are in good agreement with the design
Design of Building Foundations (DBF), AIJ [1].
curves.
(1)
K c =kh1 B(MN/m 2 )−3

kh1 = 80 E0 B 4 (MN/m3 )

361

(2)

Where, k h1 is calculated with deformation modulus,
E 0 and diameter of piles in reference. In Eq. (2), the unitof the diameter must be “cm”. Table 1 shows therelationship between the K c and E 0.
(4) The distribution of the subgrade reaction is assumed tobe the unified and triangular distributions in thefollowings.
(3)
K = Kc2Kc=
K(l − x)l

P Soil Reaction
Distribution

Where EI is the flexural rigidity of steel pile, K is thecoefficient of the subgrade reaction such as Eq. (3) or (4),and P and u are the vertical load and the lateral displacementof the pile. The boundary conditions are simple supports
such as assumption (2) in section 2.1. For the steel pile withthe unified distribution of the coefficient of the subgrade
reaction, the lateral deformation is expressed in the functionof sine curves in the followings.
nπ x(6)u = a sinl

Substituting Eq. (6) for Eq. (5) and use a variational
technique to Eq. (5), then the buckling load P cr is obtained as
the followings.
2π 
Pcr =   EIα
l 

(7)

Where α is the ratio of the buckling load of steel pile
restraint by soil to Euler buckling load in the following.
(8)K

α= n 2 +   2 c
 π  n EI

Where n is number of the buckling mode, and the number
must be selected for minimum value of α.
For the steel pile with the triangular distribution of thecoefficient of the subgrade reaction, the lateral deformation
is expressed in the function of sine curves in thefollowings.
πx
2π x
(9)
u a sin=+bll

Substituting Eq. (6) for Eq. (9) and use a variational
technique to Eq. (5), then the buckling load P cr is obtained
as the followings.
2
2
5π 
5 l 
Pcr=
  EI +   K c(10)2 l 
8π 
9  π  l  K c  256 K  l   EI −  − 4  l 
 π  4  81 l  π 
2−Table 1 Value of Kc and E0

2

2
c
4

P

K(x)=Kc

P

K(x)=2Kc

E0 (MN/m )

0.01
0.1
1

0.0047
0.047
0.47

x
l

Fig. 3 distribution of Subgrade Reaction
reaction and slenderness ratio, λ. The parameters are theshapes of subgrade reaction distribution such as unified or
triangle ones, the coefficient of the subgrade reaction such as
1 MN/m2, 0.1 MN/m2 and 0.01 MN/m2. The dots are thenumerical analyses results and the thin curves are drawn
from Eqs. (6), (7) and (10).
The bucking stresses of piles with unified subgrade
reaction distribution calculated from Eqs. (6) and (7) arealmost equal to the results of numerical analyses results on
this diagram. In general, the larger the slenderness ratio is,the buckling stress does not increase. So that the closed form
solutions for the buckling stresses of piles are suggested in thefollowing.
(11)
Pcr = 2 EIK c (l ≥ π 4 EI / K c )Eq. (11) represents the stationary value of buckling stressof Eq. (7), and thick broken lines are drawn from Eq. (11).
These lines are lower limit of Eq. (7), and are similar tothe square dots of numerical analyses results. Eq. (11) istransformed into Eq. (7), α can represent in the following.

σcr (kN/mm2)
15

Elastic Eigen Value
Unified Distribution
Triangle Distribution
Eqs.(6) and (7) (Unified Distribution)
Eq.(9) (Triangle Distribution)
Eq.s(6), (7) and (10)Smaller value from Eq.(6), (7) and (11), (12)10
1.0
Kc=1MN/m2

8

Eq. (10) can be applied in less than the local minimum
value of buckling load, because Eq. (9) represents only the
first buckling mode.

3

K c(MN/m )In case of B=50cm

(4)

2 ∫0

4

Distribution

2

2.2 Development of Elastic Buckling Load for Steel Piles
When a steel pile in liquefied soil is laterally buckled, thepotential energy U is expressed as the following [2], [3].
1 l
(5)
=
U
( EIu "2 − Pu '2 + Ku 2 )dxl PP Soil Reaction

K =0.1MN/m2c0.5
5
Kc=0.01MN/m2

2.3 Elastic Buckling Stress for Steel Piles
Fig. 4 shows the relationship between the elastic
buckling stress, σ cr for steel piles with the subgrade0Euler Buckling

50

100

150 λ (=l/i)

Fig.4 Elastic Buckling Stress for Steel Piles
362

l 

α = 2 
π 

2

(12)

Kc
EI

On the other hand, the bucking stresses ofpiles with triangle subgrade reaction
distribution calculated from Eq.(10) are almost
equal to the numerical analyses results in thesmall slenderness ratio, but the numericalanalyses results are lower than those from
Eq.(10) in the large slenderness ratio, because
Eq. (10) can be applied to less than the local
minimum value of buckling load. Then it isdifficult to apply Eq.(10) to the buckling stressesof the piles with large slenderness ratio.
As with the unified subgrade reaction
distribution, the closed form solution for the piles
with triangle subgrade reaction distribution issuggested in the following.
3 l 

α = 1+  
2π 

2

Kc
EI

(l ≥ π 4 2 EI / K c )σ5 (kN/mm2)
σy

0.4

ε=

0.3
0.2
0.1
0

Est

 σ 
+0.002   , m=17
σ 
 y
0.4
σy(kN/mm2)

σyp1=kh1By

2

E

E(kN/mm2)

206

2

4.2

1

2

3

4

kh = kh1 ( y / y1 ) −0.5

1
0

5
ε(%)

(DBF, AIJ )

0

1

2

3

4

y/y51

Fig.5 Stress – Strain Curve of Steel Piles Fig.6 Subgrade Reaction Curveu1 = a sin

P

πx

Elasto-Plastic Strut

l
l
a=
5000

(13)

The thick solid curves from Eq. (13) are fitting
well as the numerical analyses results.

m

Est(kN/mm )

E
0

p/p1
3

u2 = b sin
b=

2π x
l

l
10000

x

3. ELASTO-PLASTICBUCKLING
BEHAVIOR FOR STEEL PILE IN
LIQUEFIED SOIL

y

K=Kc
P

3.1 Summary of Elasto-Plastic Analyses
Fig.7 Initial Imperfection of Steel Piles Fig.8 Analyses Model of Steel
ModelsPile with Subgrade Reaction
In this section, FEM analyses are performed
to investigate the elasto-plastic buckling
behavior for steel piles with subgrade reaction. Material
I. Triangle Subgrade Reaction Distribution, Elastic Soilproperties of steel piles is shown as Fig.5, and then Yong’s
II. Triangle Subgrade Reaction Distribution, Elastio-Plastic Soil
III. Unified Subgrade Reaction Distribution, Elastic Soilmodulus E, yield stress σ y and strain hardening gradient E st
IV. Unified Subgrade Reaction Distribution, Elastio-Plastic Soil
2
2
for piles are E= 206 (KN/mm ), σ y =400 (N/mm ) andσ(kN/mm2)
5x
E st =4.2(KN/mm2), respectively in reference to [4]. Fig.6
λ=115, Kc=0.1MN/m2
shows the subgrade reaction curve as shown in reference [1].
4
Fig.7 shows the initial imperfection of the steel pile, andthen the initial imperfections are described as the function of3sine curves in the followings.
2π x
πx
(14)
u a sin=+ b sin
2

l
l
l
l
=
a =
, b
5000
10000

1

Fig.8 is the analysis model of steel pile with subgrade
reaction. The analysis model and boundary conditions aresame as those in the previous section. Soil is assumed torestrain the lateral deformation of steel piles, and is replaced
to elasto-plastic strut with the subgrade reaction curve in
Fig.6.
Fig.10 shows the elasto-plastic buckling behavior for steel
piles with the subgrade reaction. Figs. 10(a)~(d) represent
the compression stress, σ x (=P/A), the lateral displacement,
u/l, lateral distribution at maximum load, u/l, and flexural
moment at maximum load, M, respectively. The steel pilehas 115 of slenderness ratio, and the soil has 0.1 of thecoefficient of the subgrade reaction. The parameter is soil
conditions such as elastic or elasto-plastic property, and

0

0

0.1

0.2

εx0.4
(%)

0.3

(a) Compression Load
u/l
0.04
0.03
0.02

II IV

I

0.01
0
0

III

0.1

0.2

0.3

0.4 εx(%)(b) Lateral Displacement on Pile Top363

unified or triangle subgrade reaction distribution. Case I is
x/l
x/l
Mp
elastic soil with triangle subgrade reaction distribution, Case00
II is elasto-plastic soil with triangle subgrade reaction
My
I
distribution, Case III is elastic soil with unified subgrade
I
0.2
0.2
reaction distribution, and Case IV is elasto-plastic soil withthe unified subgrade reaction distribution.
II
III
III
0.4 IIIn Fig. 10(a), the compression load for the piles of Case II
0.4
IV
IV
and Case IV decreases rapidly after the maximum load,
because the subgrade stiffness decreases as the lateral
0.6
0.6
displacement increases. On the other hand, the maximum
load for Case I and III is about 1.5 times than that for Case II
0.8
0.8
and Case IV, because subgrade stiffness is elastic. The
maximum lateral displacement occurs at the top of pile, andfor triangle subgrade reaction distribution, the value of1
1-0.01 0 0.01 u/l-5000 0 5000 M
subgrade reaction at the top of pile is 0. So that the strength(c) Lateral Distribution (d) Flexural Moment Distribution
for Case III is higher than that for Case I.
Fig.10 Buckling Behavior of Steel Piles
In Fig. 10(b), the lateral displacements at the top of pile
for Case II and Case IV increase more than those for Case I
cannot carry the vertical load as soon as the lateraland Case III.
deformation increases, but that for elastic soil can carry the
In Fig. 10(c), the maximum deformation for Case III islarger than the others at the maximum load. It means that thevertical load after lateral deformation of the pile occurs.
pile for Case III can carry the compression load even though
The lateral displacement at the buckling of the piles isthe large lateral deformation occurs, because the soil withassumed to be y/y 1 =0.002l as shown in Fig. 10 (b) and (c),elastic subgrade and unified reaction distribution canand then the coefficient of the subgrade reaction at
restraint the lateral deformation of the pile.
y/y 1 =0.002l~0.004l is used as K c ’ instead of K c y/y 1 =1. The
In Fig. 10(d), the maximum flexural moment for Case III
reduction factor, χ is approximated as the ratio of K c ’ to K c in
becomes larger than the others, and reaches the yield
the following
moment at the maximum compression load.
(15)
=
χ =
l0 / l , l0 500cm
4. ESTIMATION OF ELASTO-PLASTIC BUCKLINGSTRESS FOR STEEL PILE IN LIQUEFIED SOIL

Table 2 shows the relationship between the length of thepiles and the reduction factor. As the length of the pile islonger, the coefficient of the subgrade reaction decreases.

4.1 Comparison between Buckling Stress for Steel Piles
with Various Soil Conditions
Fig. 11 compares the elasto-plastic bucking stress forsteel piles with various soil conditions. The vertical and: Unified Subgrade Reaction Distribution
σ /σps cr y: Triangle Subgrade Reaction Distribution
horizontal axes are the bucking stresses of the piles with1
elasto-plastic soil and elastic soil, respectively. Then they
are divided by the yield stress of the piles.
White circle represents the result for distribution of
0.75
unified subgrade reaction, black one represents the result for
distribution of triangle subgrade reaction. In the range ofless than 0.3 of es σ cr /σ y , the piles reach the maximum load
0.5
as soon as the piles deforms the laterally, so that thebuckling stresses at the maximum loads are almost equal,
0.25
regardless of the difference of soil condition. In the range ofmore than 0.75 of es σ cr /σ y , the piles reach almost yieldstress, so that the piles deform slowly after the maximum
0load. In these range, the buckling stresses for elastic and
0
0.25
0.5
0.75
1 esσcr/σyelasto-plastic soil are almost equal.
Fig.11 Comparison of Buckling Stress for Piles with Various
On the other hand, in about 0.5 of es σ cr /σ y , the buckling
Soil Conditions
stress for elasto-plastic soil is lower than that for elastic soil.
In this range, the soil restrains the lateral
Table 2 Relationship between length of piles and reduction factor
deformation of the piles until the pile buckles,
500
1000 1500 2000 2500 3000 4000l (cm)and then the subgrade stiffness for elasto-plastic
y /y 1 at =K c'
1.0
2.0
3.0
4.0
5.0
6.0
8.0
soil decreases even though that for elastic soil is
1.000
0.707
0.577
0.500
0.447
0.408
0.354
χ (=K c'/K c)constant. So that the pile with elasto-plastic soil364

4.2 Estimation of Buckling Stress for Steel
Piles in Liquefied Soil
Fig. 12 shows the elasto-plastic bucking
stress for steel piles in liquefied conditions.
The vertical axis represents the buckling
stress for steel piles, and the horizontal axis
represents the equivalent slenderness ratio.
The equivalent slenderness ratio is shown as
the following in Recommendation for Limit
State Design of Steel Structures in AIJ (LSD)[5].
λc = Py / Pcr: LSD, AIJ: DSS, AIJ: No Subgrade Reaction: Elastio-Plastic Unified Subgrade Reaction Distribution: Elastio-Plastic Triangle Subgrade Reaction Distribution: Elastic Unified Subgrade Reaction Distribution: Elastic Triangle Subgrade Reaction Distribution

σcr/σy1.25
1Pc /Py =1.0

Pc /Py =1-0.4(

λ c2 2)e λc

Pc /Py =1-0.5

λ c − p λce0.75

λc − p λc
Pc /Py =(16)

1
1.2λc2
Pc /Py =

0.5

1
λ2c

Where P y is the yield load and P cr is
Euler’s buckling load in reference [5]. In this0.25
paper, when steel piles are continuously
braced by the soil, Euler’s buckling load in
λ (=0.15)
λ ( = 1/ 0.6)p c
e c
Eq. (16) is replaced on the buckling load
0
0
0.5
1
1.5
2 λc
obtained from Eqs. (7), (8), (10)~(13). Then
this equivalent slenderness ratio is defined as
Fig.12 Buckling Stress for Piles with Subgrade Reaction
the modified equivalent slenderness ratio. The
dots are numerical analyses results, and thesecurves are drawn from AIJ codes such as LSD [5] and Design 6 ACKNOWLEDGMENT
Standard for Steel Structures (DSS) [6].
The author acknowledges support from Disaster
The results of numerical analyses are exceeded for thePrevention
Research Institute Kyoto University (DPRI), thecurves drawn from LSD and distributed as the upper-boundDisasterPrevention
Research Institute’s Joint Usage/Joint
for the curves drawn from DSS.
Research.
As the results, the elasto-plastic buckling stress for the steel
piles in liquefied soil can be approximately evaluated with theequivalent slenderness ratio by LSD.
7 REFERENCES
5 CONCLUSION
This paper evaluated the relationship between the bucklingload for steel piles in liquefied soil. It was shown that:
1) In the range of small slender ness ratio, the flexural elastic
buckling load for steel piles in liquefied soil is evaluatedfrom Eqs. (7), (8), (10) developed by the energy method.
2) The stationary value of buckling stress for the piles withthe unified subgrade reaction distribution can becalculated from Eq. (11), so that the minimum buckling
stress in the range of large slenderness ratio can be
estimated by Eq. (11).
3) In the range of large slenderness ratio, the buckling stressfor the piles with the triangle subgrade reaction
distribution can be approximated from Eqs. (12), (13).
4) From elasto-plastic large deformation analyses, in therange of less than 0.3 or more than 0.75 of the ratio of thebucking stress to the yield stress, the buckling stresses forelastic and elasto-plastic soil are almost equal, and about
0.5 of the ratio, the buckling stresses for elastic soil ishigher than those for elato-plastic soil.
5) The elasto-plastic buckling stress for steel piles inliquefied soil can be estimated by the equations in theJapanese design codes using the modified equivalent
slenderness ratio.

[1] Architectural Institute of Japan, Recommendation for Designof Building Foundations, 2001, (in Japanese)[2] Hans H. Bleich: BUCKLING STRENGTH OF METALSTRUCTURES, McGRAW-HILL BOOK COMPANY,
1952[3] Kimura Y, Tokimatsu K, “Buckling Stress of Steel Pile
with Vertical Load in Liquefied Soil”, Jour. of Structural
and Construction Engineering, No.595, Sep. 2005,
pp.73-79 (in Japanese)[4] Kimura Y, Ogawa T and Saeki E.,” The Local Buckling
Behavior of Cold-Formed Circular Tubes by Different
Manufacturing Process, Jour. of Steel Construction
Engineering, Vol.8, No.29, March 2001, pp.27-34 (inJapanese)[5] Architectural Institute of Japan, Design Standard for Steel
Structures, 2005 (in Japanese)[6] Architectural Institute of Japan, Recommendation for Limit
State Design of Steel Structures, 1998, (in Japanese)

365

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Two-dimensional finite difference pseudo-static deep-seated slope stability
analysis of embankments over stone column-improved soft clay
Farshad Fayyaz Jahani 1, Mohammad Sharifipour 2
Department of Civil Eng., Univ. of Razi, Kermanshah, IranP

P

P

ABSTRACT
A two-dimensional (2D) finite difference method was performed in this study to evaluate the factor of safety (FS)against deep-seated failure of embankments over stone column-improved soft clay. The parameters influencing the
FS against deep-seated failure of embankments over stone column-improved soft clay were investigated including the
spacing, thickness, height and friction angle of stone columns, cohesion of soft clay, friction angle, and height ofembankment fill, and existence of ground water. In this study, effort to investigate the influence of these factors withdifferent values of seismic coefficient (0.2g-0.25g-0.3g-0.35g), correspond to Iranian seismic provision (2800). The
results of these analyses are summarized into a series of design charts, which can be used in engineering practice.
Keywords: factor of safety, seismic coefficient, deep-seated failure, finite difference method, stone column.

1. INTRODUCTION
Problems of slope instability present design and researchchallenges to geotechnical engineers. Slope stability analysis
can be carried out by the limit equilibrium method (LEM),
the limit analysis method (LAM), the finite element method(FEM), and the finite difference method (FDM) [1].
In recent years, finite difference method has been widelyused for analyzing slope stability including the computationof its factor of safety (FS) [2]-[3]-[4]-[5].
Han et al. (2002) used the same finite software (FLAC) toobtain the identical difference corresponding FS values of
unreinforced and geosynthetic-reinforced slopes as the
Bishop's simplified method. Han and Leshchinsky (2004)
obtained similar results for mechanically stabilized earth(MSE) walls using the finite difference method and Bishop's
simplified method incorporated in the ReSSA software.
Also, several research exist only about evaluate the factor ofsafety of embankments in static mode , evaluate the factor ofsafety (FS) less attention has been in pseudo-static anddynamic analysis. Among the methods The finite difference
method is perhaps one of the oldest numerical techniques
used for solving sets of differential equations. In the finite
difference method, every derivative in the set of governing
equations is replaced directly by an algebraic expression
written in terms of the field variables at discrete points inspace; these variables are undefined within elements [6]. As
compared with limit equilibrium methods, finite difference
methods have the following advantages for calculating thefactor of safety of slope stability [7]: (1) no need to define a
range of trial surfaces and possible failure modes or criticalslip zones determined from the numerical results (e.g., strain
rate, plasticity); (2) no need to assume any functions forinter-slice force; (3) different structural elements used

(e.g., rock bolt, soil nail or geogrid) instead of equivalent
forces; and (5) the solution consisting of kinematicallyfeasible mechanisms.
The slope instability of embankments may develop locally,near the facing, within the embankment, or through thefoundation soil as local, surficial, general, or deep-seated
failure. Other conditions of deep-seated slope failure are
creation of Rip Rap On the downstream slope. The deepseated slope failure is also referred to as a global slope
failure, mainly induced by a weak foundation existing underthe embankment. Local and surficial failures develop at ashallow depth (mostly less than 1.2 m) due to low
overburden stress, low density, low strength, and seepage
force when the slope becomes saturated after rain. The
general slope failure typically occurs through the toe of theslope [2].
A number of ground improvement techniques have beensuccessfully adopted to prevent deep-seated slope failure,
such as sand compaction piles, stone columns, and deep
mixed columns. Stone columns have been commonly used
as an alternative to solve deep-seated slope stability
problems [8]. Stone columns can be installed using a wet or
dry method. The wet method employs a vibrating probe withjetting water to form holes to be backfilled with stones froma ground surface while the dry method uses a vibratingprobe with jetting air down to a depth and a feed pipe tosupply stone to the bottom of the probe. The stone columns
improve the ground mainly due to their higher strength andstiffness compared to the soil.
Christoulas et al. (1997) investigated the stability ofembankments over stone columns using a limit equilibrium
method with a slip circle, in which individual stone column
and equivalent area models were analyzed. They concluded
that the computed factor of safety from the individual366

column method was greater than that from the equivalentarea method. This paper presents a series of two-dimensional(2D) finite difference analyses to investigate the factors
influencing the FS against deep-seated failure ofembankments over stone column-improved soft clay inpseudo-static analysis. The finite difference method
incorporated in the software – FLAC/Slope Version 5.0 is
designed specifically to perform pseudo-static factor ofsafety calculations for slope stability analysis. The results ofthese analyses are summarized into a series of design charts,
which can be used in engineering practice.
Fig [1]. Finite difference meshes of baseline model

2. MODEL GEOMETRY
The model is consists of an embankment supported by stone
columns in soft clay under a two dimensional (2D) planestrain condition. The foundation soil consists of 9 m thickclay overlying 3 m thick sand. Due to the symmetry of themodel, half of the cross-section was analyzed using thissoftware. The stone columns were modeled as continuous
walls parallel to the centerline of the embankment. The
dimensions were selected based on a common practice in thefield. These parameters of the baseline case are provided intable [1]. The groundwater table was at the ground surface.
An “Exclude” function included in the FLAC/Slope software
was adopted to prevent the potential failure surface fromentering a 0.4 m thick surficial soil layer on the slope. In realcases the potential failure surface is prevented by
Implementation Rip-Rap On the downstream slope.

Fig [2]. Cross section of baseline model for the finite difference analysis

Table[1] General information of model used in finite difference analysis
Thickness of
Height of stone
Width of fill
Height of fillstone columns
columns
0.8 m
9m
10 m
4m
Slope of fill
Center to center distance of stone
Water level
column1V:2.5 H
3m
12 m

Fig [3] The Mohr–Coulomb yield surface in principal stress space (c=0)

3. MODEL SIZE AND BOUNDARY CONDITIONIn this program the factor of safety is strongly depended on
the mesh sizes, therefore the sizes of the model and meshwere determined on the basis of several trials, during which
the mesh was progressively refined to 0.5 m and its
horizontal boundary was extended such that zones did not
influence the development of the failure surface. The size ofthe mesh is presented in Fig [1]. In the case of boundary
condition, the bottom of model fix in two directions andnodes on the two vertical boundaries were fixed against
horizontal movement but allowed to move freely in thevertical direction as shown in Fig [2].
4. MATERIAL MODEL AND PARAMETERS
The embankment fill, the foundation soils, and the stone
columns were modeled as linearly elastic-perfectly plastic
materials with Mohr– Coulomb failure criteria as shown in
Fig [3]. The elastic properties have an insignificant effect on

the factor of safety calculation and, therefore, theseproperties are not required by the FLAC/Slope software(Dawson and Roth, 1999). Since soft clay is mostly normallyor under-consolidated, it is more critical for an embankment
over the soft clay under an undrained condition than undera drained condition [8]. Therefore, undrained cohesion was
assumed for the soft clay in this study. The installation ofstone columns may change the properties of the soft clay;
however, field studies have shown that such propertychanges are minimum, especially when a wet method is used
to install stone columns. Therefore, the change of theproperties of soft clay is ignored in this study. Consideringstone columns are formed by a cohesionless material, they
have zero cohesion and their shear resistance is influenced
by pore water pressure. The effect of the ground water on
the stability was evaluated in this study. The FLAC software
has the option to activate or de-activate the water table.
When the water table option is not selected, only theunsaturated unit weight is assigned. When the water tableoption is selected, however, either unsaturated unit weightand porosity or saturated unit weight must be input.

367

The saturated unit weight of soil can be calculated as follows[6]:
γ sat = γ unsat + n γw

(1)

In which γ sat , γ unsat , and γ w are saturated, unsaturated
soil, water unit weights, respectively and 'n' is the porosity ofsoil. In the case with a ground water table, the soil above thewater table was assigned an unsaturated unit weight. The
material properties used in the baseline case are shown in
Table [2]. Pseudo-static analysis in FLAC/Slope software isdone by introducing inclined angle and resultant seismic
coefficient in correspondence to different seismic coefficient(0.2g-0.25g-0.3g-0.35g) which exists in Iranian seismic
provision (2800).
parameter

Table [2] Material properties of the baseline model.

Thickness
Cohesion
Friction angle
Unsaturated
Unit
weightPorosity

unit

fill

Soft
clay

Stone
column

sand

m
kPa
o

4
0
34

9
25
0

9
0
38

3
0
30

kN/m3

17

16

17

19

-

0.25

0.25

0.25

0.25

The parameters listed in abstract influencing the factor ofsafety against deep seated failure of embankments over stone
column-improved soft clay in the pseudo-static analysis were
investigated in this study. In the first stage of analysis weshow that critical slip surface (maximum shear strain rates) in
pseudo-static mode for baseline model in condition of
without water and seismic coefficient 0.2 as shown in Fig [6].

Table [3] Input parameter in FLAC/Slope for pseudo-static analysis.
seismic
seismic
seismic
seismic
Parameter
coefficient
coefficient
coefficient
coefficient(0.2)(0.25)(0.3)(0.35)
resultantacceleration(g)
Inclinedangle(degree)

1.02

1.03071.044

1.059511.31

14.0362

16.67

19.29

Fig [5]. Flowchart of calculating the factor of safety in FLAC/slope

5. CALCULATION OF SAFETY FACYTOR
In the finite difference program, a shear strength reduction
technique was adopted to solve for the factor of safety (FS)value of slope stability. This program uses the shear strength
reduction technique in the calculation of safety factor. In thistechnique a series of trial FS values are used to adjust thecohesion, c, and the friction angle, ϕ, of soil as follows [6]:c trial = c / FS trial
ϕ trial = arctan (tanϕ /FS trial )

(2)
(3)

In the Fig [5] for much understanding, the stages forcalculation of safety of factor in FLAC/slope are shown in
the following flowchart.
6. DISCUSION AND RESULTS

Fig [6]. Critical slip surface without considering the water table inpseudo-static mode (shaded region).

6.1 The influence of the friction angle of stone columns
The influence of the friction angle of the stone column
material on the factor of safety of the embankment overstone column-improved soft clay for different seismic
coefficient (0.2g-0.25g-0.3g-0.35g) and different friction
angle (24-29-34-39) degree with or without water is shownin Fig [7]. The friction angles are selected based on real andapplicable project. The results show that the increase offriction angle of stone column material yielded a higher
factor of safety for the embankment system in all seismic
coefficients. As it is apparent, with the increase in seismic368

coefficient the factor of safety decreases considerably.
Furthermore, since stone column was modeled as acohesionless material, which only has a friction angle theconsideration of the ground water resulted in lower FS
values than that without a water table, because waterpressure reduced the shear strength of the stone columns.
6.2 The influence of the spacing of stone columns
The influence of the spacing of stone columns on factor ofsafety of the embankment over stone column-improved soft
clay for different seismic coefficient (0.2g-0.25g-0.3g-0.35g)
and different Spacing of stone column (3-6-9-12) meterswith or without water is shown in figure 8. The
consideration of the water table resulted in lower FS values
than that without a water table. As shown in fig [8] the factorof safety in distance between 3-6 meter higher than the otherintervals have been changes. Indicates that a proper distanceand optimal deployment of stone columns. Since stone
columns as the symmetric model since the number of stone
columns at intervals 6 and 9 meters is the same. The large
reduction in factor of safety was not expected. The factor ofsafety is increased in height of 9 meters.

The influence of the spacing of stone columns on factor ofsafety of the embankment over stone column-improved soft
clay for different seismic coefficient (0.2g-0.25g-0.3g-0.35g)
and different the height of fill (2-4-6-8-10) meter with orwithout water is shown in fig [9]. Obviously, because ofincreasing slope and weight of fill with increasing height offill factor of safety is reduced. It is clear that the factor ofsafety decreased with an increase of the seismic coefficientand height of the embankment. Height of fill will have agreat impact on Safety factor.

Fig [9]. Influence of height of fill with different (sc)

6.4 The influence of the cohesion of soft clay
The influence of the cohesion of soft clay on the safety
factor of the embankment over stone column-improved soft
clay for different seismic coefficient (0.2g-0.25g-0.3g-0.35g)
and different cohesion of soft clay (15-20-25-30-35)
kilopascal with or without water is shown in fig [10]. An
increase of the cohesion of the soft clay increased the FS
value of the fill over stone columns for all seismic
coefficients. Considering Figure effect of the cohesion
parameter is greater than the other parameters is discussed.
Fig [7]. Influence of friction angle of stone column with different (sc)Fig [10]. Influence of cohesion of soft clay with different (sc)Fig [8]. Influence of spacing of stone columns with different (sc)

6.3 The influence of the height of fill

6.5 The influence of the friction angle of fill
The influence of the friction angle of fill on factor of safetyof the embankment over stone column-improved soft clay
for different seismic coefficient (0.2g-0.25g-0.3g-0.35g) anddifferent the friction angle of fill (24-29-34-39-44) degree369with or without water is shown in fig [11]. It is shown thatthe benefit of the friction angle of the embankment fill
became less significant when the friction angle was higher
than 29°, because the slip surface developed was deeper in
the improved foundation and the contribution of thefoundation became more important. Therefore, foundationfill ratio of stiffness in applicable cases should beconsidered.
4.5 The influence of the height of stone columns
The influence of the friction angle of fill on factor of safetyof the embankment over stone column-improved soft clay
for different seismic coefficient (0.2g-0.25g-0.3g-0.35g) anddifferent the height of stone columns (3-5-7-9) meter with orwithout water is shown in fig [12]. As shown in fig [11],almost at the height values from 3 to 7 meters, because no
pass of failure curve through stone columns is seen littlechange in factor of safety occurs. In other words, failure
curve pass under of stone columns and therefore
participation stone columns in factor of safety will beignored. Also it is clear that the factor of safety decreases
with an increase in the seismic coefficient. Consideration of
the ground water resulted in lower FS values than thatwithout a water table, because water pressure reduced theshear strength of the stone columns as shown in fig [12]. Fig[12] indicates a proper height and optimal deployment ofstone columns in increasing factor of safety can be used.

Fig [11]. Influence of height of stone columns with different (sc)

6.8 The influence of the thickness of stone columns
The influence of the thickness of stone columns on factor ofsafety of the fill over stone column-improved soft clay for
different seismic coefficient (0.2g-0.25g-0.3g-0.35g) anddifferent thickness of stone columns (0.4-0.6-0.8-1.2) meterwith or without water is shown in fig [13]. Such as previous
items the increase of thickness of stone columns will result
increase of safety of factor.

Fig [13]. Influence of thickness of stone columns with different (sc)

7 CONCLUSION
A two-dimensional (2D) finite difference method was
performing in this study to evaluate the factor of safety (FS)against deep-seated failure of embankments over stone
column-improved soft clay with different values of seismic
coefficient (0.2g-0.25g-0.3g-0.35g) correspond to Iranianseismic provision (2800). Based on the numerical results, thefollowing conclusions can be drawn:
1. The results of these analyses are summarized into a seriesof design charts, which can be used in engineering practice.
2. In all parameter discussed here the consideration ofground water resulted in lower FS values than that without
ground water because pore water pressure reduced thefrictional shear strength of the improved foundation.
3. In all parameter discussed here the factor of safety
decreased with an increase of the seismic coefficient.
4. The increase in soil resistance parameters will not always
lead to an increase in factor of safety. For example, with theincrease in foundation resistance parameters the critical slip
surface is formed within the fill and will not cut the stone
columns. Therefore, foundation-fill ratio of stiffness inapplicable cases should be considered.
5. According to the charts obtained, the effect of cohesion ofsoft clay on the increase of safety factor is much more thanother parameters which discussed in this study.

8 8 REFERENCES[1] Han, J., Leshchinsky, D. “Stability analyses ofgeosynthetic-reinforced earth structures using limit
equilibrium and numerical methods,” in Proc. 8nd Int.
Conf. on GEOSYNTHETICS, 2006, pp. 1347–1350.

Fig [12]. Influence of friction angle of fill with different (sc)370

[2] Han, J., Sheth, A.R., Porbaha, A., Shen, S.L, “Numericalanalysis of embankment stability over deep mixed
foundations,” GEOTRANS, 2004, pp. 1385–1394.
[3] Han, J., Parsons, R.J., Sheth, A.R., Huang, J., “Factors ofsafety against deep-seated failure of embankments overdeep mixed columns,” Proceedings of Deep Mixing
Conference, Vol. 1.2, 2005 , pp. 231–236.
[4]

Cheng, Y.M., Lansivaara, T., Wei, W.B.,“Two-dimensional slope stability analysis by limit
equilibrium and strength reduction methods,” J. of
Computers and Geotechnics,vol 34, 2007, pp. 137–150.

[5]Han, J., Hong, Z.S., Shen, S.L, “ Stability of levees oversoft soil improved by deep mixing technology,”
GEOCONGRESS ASCE ,2008, pp. 716–723.
[6] Itasca Consulting Group, Inc., 2006. FLAC/Slope User's
Guide, Version 5.0. 84.
[7] Cala, M., Flisiak, J.,( 2001), “Slope stability analysis
with FLAC and limit equilibrium methods,” In: Bilaux,
Rachez, Detournay, Hart (Eds.), FLAC and Numerical
Modelling in Geomechanics. A.A.
BalkemaPublishers, pp. 111–114.
[8] Cooper, M.R., Rose, A.N, “Stone column support for an
embankment on deep alluvial soils,”. Proceedings of theInstitution of Civil Engineers, Geotechnical Engineering
vol 37, 1999, pp. 15–25.

371

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Parametric Study of Geo-grid Soil Reinforced Retaining Walls under
Dynamic Loading by ABAQUS Software
Prof.Atrchian Mohammad Reza1, Prof.Daghigh Younes2 Mr.Ghanbari Namin Hamed 3
Islamic Azad University of Zanjan, Iran
ABSTRACT
Observation of reinforced soil retaining walls with geosynthetics system, after earthquake and labratory studies
indicate the suitable inflection behavior of these walls. For better underestanding of these kind of structures
under dynamic loads, in this research, at frist we use finite element method (FEM) software (Abaqus) for
numerical modeling of reinforced soil retaining walls under eathquake. At this phase, in order for testing andcomparison of the software,results of a similar sample with small scale in shake table test ,which was
impelemented in 2005, is compared with the results of test and analysis by the software. After correcting themodel and making sure of the software's accuracy, models of several reinforced soil retaining walls withgeosynthetics of 10 meter height and variable length and distance of geogrid layers are provided and dynamic
analysis is performed on them. Harmonic equivalent load is used for earthquake acceleration in dynamic
analysis. After impelementing the analysis, by concentrating on the forces in the geosynthetics,diagrams ofvertical variable displacement on soil and wall are prepared that include soil subsidence, horizontal
displacement or walls bulging. Research is implemented parametrically and by studying the effects of length
increase and geogrid layers spacing on displacement and results are obtained on this basis.
Keywords: Retaining walls; Reinforced soil; Earthquake; Geo-grid; Finite element method (FEM).


1. INTRODUCTION
Reinforced soil retaining walls on land areas that have nonstable and weak slopes with weak beding makes it
possible to build stable and flexible structures against shaking loads(earthquake) with increasing load bearing
capacity. This results in decrease of the structures subsidence, falling, creep and swelling capability and increase ofsafety factor and Improvement of shear strength parameters of soil[1]. Soil has a good resistance against stress andwith reinforcing Geogrid that have a high tensile capacity with facing as a protective soil between Geogrid layers, it
results in a stable system. Friction between soil particles and Geogrid results in a cohesion strength that has a direct
relationship with Geogrid tensile resistance[2].
Points mentioned below are considered as the most important criteria to choose components of reinforced soil
retaining walls:
1.1.Criteria to choose embankment: Soil that is used in this system must be contorting with the criteria and related
standard in terms of optimum moisture and aggregate gradation so that the required friction between soil andreinforcement is ensured. Two kinds of criteria for these system are proposed:a) Chemical and electrochemical criteria to prevent corrosion of the reinforcement such as limitation of chlorosulfate
purity, PH limitation, and amount of organic material in soil.

1. Professor, Islamic Azad University of Zanjan, Iran , Dr.atre@yahoo,com
2. Professor, Islamic Azad University of Karaj, Keyhan khak Co. (prt)Ltd, Daghigh_y@yahoo.com
3. Geotechnical student, Islamic Azad University of Zanjan, Iran, Hghanbarinamin@yahoo.com

372b) Geotechnical criteria such as prevention of vacuolate pressure and condensation of soil in horizontal layers.
1.2.criteria to choose the reinforcement: Geosynthetics alike artificial textiles are made from polymeric material
such as polyethylene and polypropylene and include Geotextil, Geogrid, Geomembranes, and Geocomposits and …
and are employed as reinforcement, separation, filteration, drainage and moisture barier.
In this research, the purpose of utilization is reinforcement. Therefore, use of Geogrids in which are only applied forthe purpose of reinforcement is proposed. choosing reinforcement spaces and length of them is implemented withcalculation of the inner stability of the buildings. Resistance of the reinforcement against strain and it’s flexibility
are the main factors of endurance.
1.3.criteria to choose facing: rigid facing are usually utilized for permanent reinforced soil structures for persistenceof better appear are and ease of construction. Based on laboratory and field investigations, facing rigidness in total
local state can help us in decreasing shape change while increasing structure resistance. These facing differ in termof shape, type, and profile[4] :a) Wrapped Around Facingb) Modular concrete blocks Facingc) Pre-cast concrete Panels Facingd) Gabions Facing

Fig.1: facing of shape, type, and profile

2. DYNAMIC STABILITY AND LITERATURE REVIEW
2.1 Dynamic stability
When earthquake occurs, seismic waves move from very deep depths of the earth to the surface and cause
vibrations. The intensity and durability of this vibration is related to the intensity and magnitude of the earthquake
and the environment characteristics. These waves face the decrease or increase in domain in different layers of theearth and depreciate the vibration in a specific frequency while resonate it in another frequency. This behavior is due
to different characteristics of these layers and filtering behavior of them. Generally, soil slopes fracture is due togravity effect on a mass of material that can be slid gradually and move down on an individual layer. This
movement is either minor and related to failing of a few stones, or major in a vast area and cause serious damage on
the infrastructure. Increasing the slope height by filling on the slope and cutting near the toe will increase instabilityof the slope. Owing to these changes (increasing the height and angle) thrust will increase and tangential componentof thrust (shear stress in the slip surface) in the fracture surface that is one of the instability reasons will cause themass to move down. Another reason of increasing the shear stress along the potential fracture surface is increasing373

the weight of the soil zone. This pressure increment can be due to more urban construction on the slope or due tonatural reasons such as falling upper slippy zones. In all cases shear stress will increase and soil zone will be instable
along potential surfaces. Also, making vibration either natural or artificial will increase the shear stress along thepotential fracture surfaces. During the earthquake and at the time of earth vibration, the spread waves in the soil
zone change the level of forces and cause the fracture of the soil (along the weak surfaces) that is under the cyclicload and do not resist against stresses [3].
2.2 literature review
During the past years, lots of experimental research is performed to investigate the seismic performance of these soil
walls by employing shaking tables and centrifuges. Richardson and Lee (1975) performed the first experimental
investigation in this field. They modeled reinforced soil walls and introduced the first design criteria according tolimit equilibrium method to calculate reinforced soil walls. In summary, Sakaguchi (1996), Matsu et al. (1998),
Koseki et al. (1998), Siddharthan et al. (2004) and Ling et al. (2005) investigated these walls under seismic loads.
Bathurst and Cai (2002) performed a quasi-static investigation on a wall that was reinforced with Geogrid and was
subjected to Northridge earthquake. They found that these walls have suitable safety factors in the case of various
fracture conditions. Ghalandarzadeh and Mirlatifi (2003) performed some shaking table experiments on thereinforced Geogrid walls. Then, Ghalandarzadeh and SaberMahani (2008) did a series of shaking table experiment
to investigate the mode shapes and seismic response of these walls [5].
3. MATERIAL PROPERTIES AND MODEL CONSTRUCTION IN ABAQUS SOFTWARE
One of the most important parameters in the lateral soil pressure on the reinforced soil walls is the soil shear strengthparameters. For the embankment material, Mohr-Coulomb (elastopelastic) model and drained behavioral are used
(Table 1):
Table 1. Embankment and bed soil parameters used in the modelingsoilparameters

Soil
unit
weight
above
g.w.l

Soil
unit
weight
below
g.w.l

(Kx)

(Ky)

Young'smodulus(Eref)

Poisson's
ratio
( )

Cohesion(Cref)

Frictionangle(φ)

Dilationangle(ψ)

Unifiednamemodeling zone
Embankment

17

18

1

1

2.5E+4

0.3

10

30

0

SW-SM

bed soil

19

20

0.413

0.413

3E+4

0.3

30

28

0

SC

In order to model, a wall with H=10m, θ=90 (angle with horizontal), =0.2 and continues concrete surface and a
Geogrid made of Polyethlene, covered with PVC, meshed and with an axial stiffness of 3200, 2200, 1200 KN/m areconsidered [6].
In the case of Geotechnical modeling, usually the considered zone is big enough to be assumed infinite incalculations. Existence of any virtual boundary in these problems will cause the return of introduced waves into thecalculation area after encounter to these boundaries. However, in reality, these waves can distribute in the infinite
environment and be absorbed and do not return to the calculation environment. The method that is used in the
program is introducing absorbing boundaries in the boundary reasons to almost completely absorb the entrance
volume wave energy and prevent reflection.
After modifying the model under the name of Verification and Calibration with a similar experimental model, webuild the model and perform sensitivity analysis of mesh changes and model dimensions. In the case of meshing, theaverage value is related to the condition that mesh size is fine. In the case of mode dimensions, all models are under
the gravity load in such a way that the soil is slipped and slip cycle can be observed. According to the output ofvarious models when the depth of the considered zone around the wall has a minimum depth equal to the height of374

the wall, stress and strain calculations can be performed without any limitation. Fig.2 shows the constructed modelof the reinforced Geogrid wall. In order to parametrically investigate the Geogrids’ dimensions, their length isnamed(a) and vertical distance of Geogrids is considered as (b):Fig.2: constructed model of the reinforced Geogrid wall

4.DYNAMIC ANALYSIS IN TERMS OF PARAMETRIC COMPARISON OF LENGTH AND VERTICALDISTANCE OF GEOGRID IN ABAQUSIn order to investigate the effect of reinforcement on retaining walls, several different models are introduced. To
determine the longitudinal parameter (e.g. a) the length of Geogrid is considered as 5.5 meters and is increased with
1.5 meters increments to 13 meters that is 2 meters less than supporting soil. Having 6 models, the minimum
altitudinal parameter (e.g. b) of Geogrids are considered as 0.5 meter and is increased with 0.3 meter increments to 2
meters forming 6 more models. Totally 36 models (w1-w36) are considered in this study.
In this study dynamic analysis is performed using equivalent cyclic load of each record with a period that is definedin the seed graph for different magnitude in a distance less than 40 kilometers of the source fault. Since the mostsuitable method to investigate the vibration characteristics of earthquake is to change vibration movements fromtime domain to frequency domain using Fourier series, the frequency limit of the biggest accelerogram energy canbe determined by studying the earthquake records and Fourier spectrum. Using this Information, the rate and thepattern of energy distribution in each record can be found. Also, another important application of Fourier domain
spectrum of an earthquake is determining the excited vibration modes of a structure. This spectrum is employed toanalyze the structure that is subjected to the earthquake, determine the effective modes of the structure and calculate
structural response with an acceptable accuracy. Frequencies and damping of different modes in each step can becalculated for the whole structure by subjecting a selected degree of freedom to the earthquake and finding the
response of that degree of freedom [7]. To calculate the equivalent number and amount of accelerograms
SeismoSignal software is employed where earthquake accelerogram is the input and frequency is the output.
Northridge earthquake happened in California State of USA in 1994 and caused severe damages. The acceleration ofthis earthquake in X direction is shown in fig.3 and fig.4 shows the output of SeismoSignal that is an equivalentharmonic (sinusoidal) load. In order to perform dynamic analysis and parametric investigation, the acceleration of
the harmonic load is applied to the base point of the model as a horizontal component. This acceleration is applied tothe system when it is in the static equilibrium. The considered acceleration for this wall is a wave with variable
amplitude and the mathematical equilibrium of it is presented as:(1)
Where α=5.5, β=55 and ξ=12 are constant values. f is the frequency acceleration and t is the dynamic analysis time.
The selected frequency is around the natural frequency of the system.

375

Northridge-X

0.3…Northridg

0.2
0.1
0
0.1- 0

2

4

6

8

10

12

14

16

0.20.3Fig.3- Northridge accelerogram in the X direction and in 14.96 seconds
3
2

acc

1
0
1-

0

2

4

6

8

23-

Dynamic time
Fig.4- Subjected accelerogram to the retaining wall

The accelerogram presented in fig.4 is applied to the retaining wall and the wall displacement according to its heightis the output of the models. The accelerogram is applied to the base points that are connected to the earth and the end
point that is connected to the earth is restrained in the Y direction and the accelerogram is applied in the X direction.
After applying the load to W1 model (a geogrid wall with a length of 5.5 meters and a vertical distance of 0.5
meter), as it was predicted the vibration cycle is formed on the soil and the maximum settlement is 18 cm and isshown in the red part of fig.5.

Fig.5- Total displacement contour

376

Displacement(m)As it is shown in fig.5, the soil behind the wall is vibrated and the maximum horizontal displacement isapproximately in the middle height of the wall. This will cause a deflection bluging at the center of the wall.
In order to investigate the effect of the Geogrid’s length on the wall’s stability, the models with the same distanceand variable length of Geogrid is compared. For this purpose models 1 to 6 are compared. The horizontal andvertical displacements of these models are compared and the results are presented in graphs 1 and 2 in terms of thevertical and horizontal displacements versus the height from the wall’s base.

0.2
0.18
0.16
0.14
0.12
0.1
0.08
0.06
0.04
0.02
0

W
1
W
2
W
3
W
4
0

2

4

6
8
Wall Height(m)1012Graph 1-horizontal displacement versus the height from the wall’s base for models w1 to w6

12

8

W1

6

W2
W4

4

W5
W6

Wall Height(m)

10

2

W3
0
-0.3

-0.25

-0.2

-0.15
-0.1
-0.05
0
0.05
Displacement(m)Graph 2- vertical displacement versus the height from the wall’s base for models w1 to w6

As it can be seen from graphs 1 and 2, model w1 has the biggest displacement in both vertical and horizontal
direction while w6 has the lowest displacement. W1 is moved 18 cm in horizontal and 28 cm in vertical direction
while these displacements are 9 and 12 cm for w6 that is showing about 50 and 57 percent reduction.
According to the above graphs it can be concluded that in the models with the same vertical distance (e.g. w1 tow6), the horizontal displacement’s difference is very low; however, vertical displacements are showing a big
difference. This means that the length of Geogrid has more effect on the vertical displacement. The horizontal
displacement of the wall and vertical settlements are shown in Table 2.

377

Table 2- Horizontal and vertical displacements in the investigated models
Uy(cm)
Ux(cm) model Uy(cm) Ux(cm) model Uy(cm) Ux(cm) model
37
33
W25
33
25
W13
28
18
W1
31
30
W26
28
24
W14
25
17
W2
27
28
W27
24
22
W15
21
15
W3
23
25
W28
18
19
W16
16
13
W4
22
23
W29
16
17
W17
14
11
W5
21
21
W30
15
15
W18
12
9
W6
38
37
W31
35
29
W19
31
22
W7
33
33
W32
30
27
W20
27
20
W8
29
31
W33
25
25
W21
23
18
W9
26
28
W34
20
22
W22
18
15
W10
25
27
W35
19
20
W23
15
13
W11
23
25
W36
18
18
W24
14
12
W12
Graphs 3 and 4 are presenting horizontal and vertical displacements for models w1, w7, w13, w19, w25 and w31
where Geogrid length is the same but different distances between Geogrids are considered.
12
W1

Wall Height(m)

10

W7

8

W13
6

W19

4

W25

2

W31

0
0

0.1

0.2

0.3

0.4

displacement(m)Graph 3-horizontal displacement versus the height from the wall’s base for models w1, w7, w13, w19, w25 and w31
12Wall Height(m)

10
8

W1
W7
W13

6
4
2

W19
W25
W31

0
-0.4

-0.3

-0.2
-0.1
0
0.1
Displacement(m)Graph 4- vertical displacement versus the height from the wall’s base for models w1, w7, w13, w19, w25 and w31378

As it can be seen from the mentioned models that have Geogrids of the same length and different distances ,horizontal displacements have a significant difference compared with eachother. In model w1 with Geogrid lengthof 5.5 m and vertical distance of 0.5 m the horizontal displacement is about 18 cm while in model w31 with thesame Geogrid length and vertical distance of 2 m the horizontal displacement is 37 cm that is two times (51%) morethan w1; this difference is about 10 cm (e.g. 26 %) for vertical displacement. This shows that the effect of Geogrid’s
length on the vertical displacements and the effect of Geogrids’ distances on horizontal displacement are more.
5. CONCLUSION
After applying the dynamic load to the models, the following conclusions can be drawn:[1]. The soil walls reinforced with Geogrid have special characteristics; having enough strength, they are flexible
and therefore more suitable for the regions with higher risk of earthquake and weaker bed.
[2]. By introducing some changes in the length and distance of the reinforcement layers and also by duplicating thereinforcement layers in the sections with the maximum force, an optimum design can be achieved. Using thisoptimum design the structure will be economical, more stable, have a better performance and need less constructiontime.
[3]. Frequencies and damping of different modes in each step can be calculated for the whole structure by subjecting
a selected degree of freedom to the earthquake and finding the response of that degree of freedom. The frequency of
the first 10 modes is calculated and is 3.71, 4.06 and 5.1 for the first, second and third modes.
[4]. To have an accurate investigation and analysis, first of all, a sensitive analysis is needed to be done on thedimensions and the kind of mesh size. After performing the sensitive analysis, a depth of the foundation equal to thewall’s height and a height of the soil behind the wall equal to 1.5 times of the wall’s height are found to be suitable.
[5]. The vibration cycle is formed on the back of the wall for all models; the total transformation nature is constantand in the form of vibration cycle and only the displacements amount change. The maximum horizontal
displacement is approximately in the middle height of the wall and the maximum vertical displacement (settlement)
occurs at the top of the soil behind the wall.
[6]. The change in the vertical distance between Geogrid layers changes the relative horizontal displacements andsettlements. Although increasing the vertical distance of Geogrids increases the vertical displacements, butaccording to the introduced graphs, it can be concluded that the effect of vertical Geogrids on the horizontal
displacements is significantly more. Therefore, it is better to change the distance parameter of the Geogrids tocontrol the horizontal displacement.
[7]. Although increasing the length of Geogrids is efficient to increase the horizontal displacements and have a
direct relation with it, but the effect of Geogrid layers length on the vertical displacements and soil settlements ismore. So, in cases that the soil settlement is of crucial importance, in order to control the vertical displacement of theretaining walls, the effective parameter is the length of Geogrids.

6. ACKNOWLEDGEMENT
The authors would like to express sincere gratitude to prof.daghigh younes,Keyhan Khak Co. (prt)Ltd, for thecooperation and advice.

379

7.REFERENCES[1]. Shafabakhsh, Gh., 2005. “Reinforcing embankment and the bed of the roads using Geogrids”, Education,
Research and Technology Coordination, Transportation Group, Road and Urban Development Ministry.
[2]. Vafaee, R., Aliari, H., and Hamidi, H., 2008. “Stabilizing analysis of trenches using Geogrids”, Faculty ofEngineering, Azad University of Tabriz.
[3]. Mirlatifi, A.R., Fakher, A., Ghalandarzadeh, A., 2009. “Reinforced soil walls’ deflection investigation underearthquake”, Engineering Faculties, University of Tehran.
[4]. Vidal, H. 1966 “La Therre Armee Annles Inst ”. Tech. Du. Batim. Suppl. Vol 2,No. 223-224[5]. Richardson,G.N. and Lee,K.L.(1975). “Seismic Design of Reinforced Earth Wall” , Jol. Of GeotechnicalEngineering, Vol.101, NO.2,PP.167-188.
[6]. ASTM Standard American Society For Testing and Materials, West Conshohocken ,PA,USA.
[7]. Keefer,D.K. ,Wilson,J.P. ,Youd,T.L. 1985 “Predicting Areal Limitis of Earthquake-Induced Landsliding” , U.S.
Geological Survey,Reston,Virginia,Professional Paper PP. 317-345.

380

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Sensitivity analysis in soil nail wall for Tehran-Tabriz railway byconsidering the coefficient of variation for effective parameters in stability
analysis
Mohammad Moghadaripour 1, Ardalan Akbari Hamed2, Ali Ghozat3
1. International Institute of Seismology and Earthquake Engineering, Tehran, I.R. Iran
2. Islamic Azad University, Central Tehran Branch, Tehran, I.R .Iran
3. Shiraz University, Shiraz, I.R .Iran
Abstract:
The construction operations of the underground section of the Tehran-Tabriz railroad executed by cut and covertunneling method. But Due to the far distance of piles in some parts, vertical soil nail wall was used as the
support of excavation system. This paper presents the sensitivity analysis of this supported wall. The effect of
uncertainties has been studied through the investigation of changes by means of coefficient of variation (COV)of effective parameters such as cohesion, friction angle, and specific gravity on factor of safety. The
computational results show that Some parameters are very sensitive to the changes of COV and some of them
are not. Moreover, as the data scattered is more the dispersion in factor of safety is also more.
Keywords: Sensitivity analysis, Soil nailed wall, Monte Carlo simulation , Uncertainty

1.Introduction:

installing closely spaced grouted steel bars, called“nails” into a slope or excavation as construction
proceeds from the top to down. Nails arecommonly referred to as “passive” inclusions. The
nail bars are forced into tension as the grounddeforms laterally in response to the loss of support
caused by continued excavation. The grouted nails
increase the shear strength of the overall soil massand limit displacement during and after excavation.
A structural facing connected to the nails is used
when the slope angle exceeds a predetermined
critical value or when environmental conditions
would cause deterioration of the exposed soil faceover its design life [1].
For uncertainty analysis, there should be somerequirements like data collection, model selection,assessing the uncertainty in the selected model andits significant characteristics, and a verification ofassumptions made. The evaluation of the data setsneeds recognition of the type of the uncertainties,
whether the variables are dependent or
independent, whether the observations are
independent and whether uncertainties noted are
the result of a combination of uncertainties inunderlying parameters and proceeds.
The uncertainties associated with a geotechnicalproblem can be divided into two categories:
Aleatory and epistemic. Human errors andomissions, which are not covered herein, would fall
into a third category. Within a nominally uniform

The construction operations of the undergroundsection of the Tehran-Tabriz railroad to the lengthof 4,165 meters in district 17 and 18 of Tehran
commenced on August 5th 2008 by cut and covertunneling method. This new railroad will enable
more efficient, reliable, and environmentally safe
travel.
Briefly the construction stages of project consist of
the following steps:
Step 1 : Installation of excavation support/tunnel
structural walls (pile walls)
Step 2: Construction and waterproofing of thetunnel top slab tying it to the support of excavation
walls
Step 3: Excavation of tunnel interior
Step 4: Construction of the tunnel floor slab
Due to the far distance of piles in some parts,
vertical soil nail wall was used as the support of
excavation system. The soil nail wall was regularly
shaped and reached total heights up to 10 meters.
Tight excavation requirements were defined for the
project due to the location of the building, requiring
fairly precise wall positions and shotcrete
tolerances.
Soil nailing is a technique used to reinforce andstrengthen existing slopes. Soil is a poor structural
material because it is weak in tension, whilst Steelis strong. The fundamental concept of soil nailingis that soil can be effectively reinforced by381

geological layer, the soil properties can be affectedby both aleatory and epistemic uncertainties[2]:• Alearory uncertainty represents the natural
randomness of a variable. Examples ofaleatory uncertainty are the spatial
variation of a soil parameter within anominally uniform geological layer, thevariation in the peak acceleration of adesign earthquake with a given return
period, the variation in the ocean wave
height or wind force, and so on. The
aleatory uncertainty is also called theinherent uncertainty. Aleatory uncertainty
cannot be reduced or eliminated.
• Epistemic uncertainty represents theuncertainty due to lack of knowledge on avariable epistemic uncertainty includes
measurement
uncertainty,
statisticaluncertainty (due to limited information),
and
model
uncertainty.
Statisticaluncertainty is due to limited informationsuch as limited number of observations.
Measurement uncertainty is due to for
example imperfections of an instrument orof a method to register a quantity. Model
uncertainty is due to idealizations made in
the physical formulation of the problem.
Epistemic uncertainty can be reduced, andperhaps eliminated, by collecting moredata and information, improving the
calculation methods.

1- Choosing input parameters which seem to besensitive in analysis. Soil strength parameters
including cohesion (C) and friction angle (φ),
specific gravity and seismic pseudo static
coefficient (k h ) are taken into account as input
parameters. The k h is defined as fraction of thenormalized horizontal acceleration (A m ), whichacts at the center of gravity of the wall-soil mass
(AASHTO, 1996). A m is a function of thenormalized peak ground acceleration coefficient(A), which is the actual peak ground acceleration
normalized by the acceleration of gravity (g), and isdefined as[3]:Am =A (1.45 - A)
(1)
2- The effect of uncertainties has been studied
through the investigation of changes by means ofcoefficient of variation (COV) of theaforementioned
parameters.
The
completedefinition of COV will explain in part 4 (problem
analysis). The COV of each parameter in the
problem will change while other parameters are
considered constant. For example the friction angle(φ) of the first layer is 5 degrees (see table I) andthe coefficient of variation for this parameter has
been chosen as 4, 8 and 12%(the process of
selecting these COV will explain later). For any
selected COV, 2000 random samplings werecarried out. Every sampling concluded to a special
safety factor (F.S); therefor 2000 safety factors will
be elicited. Then the COV of these safety factors
will be calculated and the trend of F.S changes isobservable for this parameter in first layer.
3- This procedure will be carried out for all
remained parameters to discern the trend of F.Schanges.
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The uncertainty sources have been introduced
above. The problem of soil nailed wall also
includes most of the mentioned uncertainties;
hence selection of the efficacious data might bedifficult and in practice the calculation will be
performed by considering the critical condition. In
order to provide an accurate modeling in slope
stability problem, exact observation must be done.
In this research for identifying the sensitive
parameters, the effect of scattering in input data inanalysis has been considered. In other words, theinfluence of scattering in input data on safety factor
will be taking into account .As the result, Sensitive
parameters and their effects in the design of soil
nail wall will be introduced.

RR

3. Tehran Tabriz soil nail wall
The soil nailed wall has a depth of 10 m. The
properties of the soil layers are presented in Table
I. The model used in analysis was Mohr–Coulomb
model. Fig. 1 illustrates a profile of the soil nailedwall.

2. Sensitivity analysis
In problems with several uncertain parameters, itwould be useful to know which parameters havetangible effect in the result of the slope stability
and which do not have. A sensitivity analysis isperformed to identify how each variable willchange the result. In this approach, typically eachvariable in the problem is changed, and then thechanges in the final result are observed.
The procedure of sensitivity analysis is as follows:

Fig. 1 Profile of soil nailed wall with depth of 10m
Table I. Geotechnical properties of excavation
Thickness
Specific
Friction Cohesionof layergravity(KN/m3)
angle(KN/m2)
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382

P

P

P

(m)
0-2
2-10

of the same parameter. A summary of the COV
values reported in the literature is presented inTable III [4].

(°)
20
20

25
29

5
30

The stability analysis was carried out using
GEOSTUDIO-W/SLOPE which is based on theMorgenstern-Price method. Recommendations forthe minimum factor of safety for pull out resistanceand nail bar tensile strength are set at 2 and 1.8
respectively [3]. Table II presents Characteristics ofnails used in the designed excavation.

Table III. The range of COV reported in literature(%) [4]
Specific friction
Soil properties Cohesion
gravityangle
COV (%)
30-45
5-15
4-12
The location of the soil nailed wall project is in
Tehran. The seismic acceleration of this region is
0.35g; so the average of pseudo static acceleration
coefficient is calculated as follows [3]:(4)Table II. Characteristics of nails used in thedesigned excavation(5)
According to the seismic hazard research report,
the coefficient of variation (COV) for the pseudo
static acceleration coefficient was considered as
25% to 35%.
Based on the sensitivity analysis steps mentioned inpart 2, the trend of changes in (COV) for F.S
against effective parameters are achieved (see fig. 2
to 4).

4. Problem analysis through considering thecoefficient of variation for effective parametersin stability analysis
In the current paper, In order to get better
calculation parameters for slope stability analysis,
as well as provide a better theoretical basis for
slope design and reinforcement, the effect of
uncertainties has been applied through theinvestigation of changes by means of coefficient ofvariation (COV) for effective parameters such as
cohesion, friction angle, and specific gravity. The(COV) represents a relative and dimensionlessmeasure of dispersion and is expressed as:

Fig. 2 Effect of friction angle variation on safety
factor(2)
Where µ is average and σ is standard deviation of
the selected parameter.
The COV has been commonly used to describe thevariation of many geotechnical soil properties andinsitu test parameters. Note that the mean, standard
deviation and COV are interdependent - knowing
any two will give the third. In practice, it becomes
convenient to estimate moments of geotechnicalsoil parameters where little data are available(sparse data) by assuming that the COV is similar
to previously measured values from other data sets383

heeding more, dispersed data lead to ascend
dispersion in the factor of safety.
- Fig.2 shows the trend of COV variation of thesafety factor along the changes of friction angle (φ)
for both layers. As it can be seen, influence of thefirst layer friction angle is negligible rather than thesecond layer. In the second layer the range of F.S
variation is between 3 to 9%.
- It is clear from Fig.3 that the effect of second
layer cohesion on safety factor is more significant
compared to the first one. This range has beenchanged from 10 to 13%. The amount of F.S
variation is high which shows the importance of
second layer cohesion on safety factor.
- As it is observable from Fig.4, the specific gravity
of both layers does not have any meaningful effecton safety factor. As an example for the second
layer the F.S changes is between 2 to 4%. This
amount of changes does not seem to be mucheffective on the results.
- From Fig.5 it can be deduced that K h has
considerable effect on safety factor. The range ofchanges for COV of F.S is between 10 to 13%. It
shows the sensitivity of the results to K h .
It can be concluded that the scattering in C and φ ofthe second layer and the K h has influential effecton the safety factor; therefore, in estimating thevalue of effective parameters, additional
consideration should be done. Moreover, inpractice the result of slope stability would be more
valid if the lower limit of sensitive parameter usedin the analysis.

Fig. 3 Effect of cohesion variation on safety factor

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Fig. 4 Effect of specific gravity variation on safety
factorR

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References[1] FHWA -SA-93-068, Federal Highway
Administration .April 1994, Soil Nailing Field
Inspectors Manual, prepared for us Department ofTransportation, pp. 1-2.
[2] D.V Griffiths, A.Fenton,(2007), Probabilistic
Methods in Geotechnical Engineering , PP 72[3] FHWA 0-IF-03-017, Federal Highway
Administration (2003), "Geo technical Engineering
CIRCULAR NO.7 Soil Nail Walls" prepared for us
Department of Transportation, pp. 4-8

Fig. 5 Effect of K h variation on safety factor
RR

5. Conclusion
The results of sensitivity analysis of soil nailed
wall in the project of Tehran-Tabriz rail road havebeen reported. Four parameters including cohesion(C) and friction angle (φ), specific gravity andseismic pseudo static coefficient (k h ) wereconsidered in the analysis. Several results andfeatures base on the analysis are summarized
below:- The COV has been put to the practical use todemonstrate the variation of soil properties andscattering of input data. As it is observed by
R[4] Allen L. Jones, Steven L. Kramer and Pedro
Arduino,2002, Pacific Earthquake Engineering
Research Center, Estimation of Uncertainty inGeotechnical Properties for Performance- Based
Earthquake
Engineering,
UniversityofWashington, pp27-29

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Evaluation of Strength Parameters of Improved Collapsible Soils with Current
Additives (Case study: Atrak zone, Iran)
Mohammad Arab Ameri1, Samaneh Kargar2 and Mehdi Arab Ameri3
1MSc, earthquake engineering, Iran2
MSc, water structures engineering, Iran
3
MSc, geotechnique engineering, Iran
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ABSTRACTThe collapsible soils show unusual behavior under specific situations such as uncommon loads and/or dynamic forces
like earthquake. It causes the decrease of volume and large settlement with failures and cracks in the supported
structures. In this paper the properties of collapsible soils and the effect of additives such as cement, lime and bentonite
on physical and strength characteristics of them is investigated. For these reason, the strength tests, such as; direct shear,unconfined compressive strength and collapse potential of soil tests have been carried out. The results show that theadditives cause improvement of soil properties with cement faster effects in comparison with the lime.
Keywords: Collapsible soils, Improvement, Cement, Lime, Bentonite

1. INTRODUCTION
Generally, all soils that during construction of new structures
or use of existing structures encounter with problems areclassified in problematic soils group. These soils exist in
many forms such as soft soils, swelling soils, absorption ofhigh water soils … [6]. Collapsible soils are classified inproblematic soils from point of view geotechnique andgeology engineering [4]. Investigation of collapsible soils
becomes to necessity regard of wide spread of these soils andnecessity of designing of dams, channels and other structuresin these zones. Geotechnical study about collapsible soils isincreasing and fundamental understanding of geotechnicalproperties and behavior of these soils is very necessary andimportant job.
In 1959, Clevenger exhibited one of the first criteria for soil
collapsibility potential [2].Gibbs and Bara (1967) carried out
researches about utilization of clay mortar and its injection toloess soils [3]. Bell (1993) suggested use of emulsion
mixtures of cement or lime for loess stabilization [1].Turanand etal. with investigation of compressing injection method
derived that effect of earthquake and its influence such asliquefaction is minimized in these improved soils [5]. Huat
and Kazemian (2009) with comparing injection method anddeep soil mixing method concluded that deep soil mixing
methods that extremely used for soil improvement are
acceptable completely.
2. General method of investigation
In this study, examinations have been carried out in field andin laboratory. Thus soil samples of this zone were sent tolaboratory for identification of geotechnical properties. Thenthe stabilization of soil was investigated with adding thevarious percentages of lime, cement and bentonite to this soil.

Therefore the tests were done on them again. In the next stage,
the results of the tests were analyzed accurately and the
optimum percentage of additives was specified finally.
3. Field program
One of the most important operations in soil studies is subsoil
examination and soil exploration. In this study, withexcavating the some boreholes, both sub layer of soil was
examined and properties of soil layers were probed withsampling of soil and implementation the filed tests. In thisproject, three boreholes in each zone (overall 10 zones) were
drilled for identification of "Atrak border line". Depth of eachborehole was five meters. The in situ density of soil was
obtained by implementation the sand-cone test (ASTM
D1556). Undisturbed and disturbed samples were taken.
Water level didn’t observe up to five meters of ground surface.
Results of the sand-cone tests showed that average of the insitu unit weight of soil was 15.3 KN/m3.
4. Laboratory tests
4.1 Physical tests- Test for determination of water content of soil samples was
carried out by ASTM D2216. The results of these tests
showed that moisture content of samples was 8% averagely.
- Tests for determination of specific gravity of soils (ASTM
D854) were done on samples and this parameter was obtained
2.64 averagely.
- Tests for particle size analysis of the soils (ASTM D422)
were carried out. The soil of Atrak zone was classified toclayey silty and sandy silty loess.
- Tests for determination of liquid limit, plastic limit andplasticity Index (ASTM D4318) were done on the samples.
The soil of the zone averagely had liquid limit up to 28% andplastic index up to 8%. It is important to say that the soils ofsome boreholes didn’t have plasticity properties.

385

- The results that obtained from standard proctor compaction
tests (ASTM D698) showed that the optimum moisture
content and the maximum dry unit weight were 14.6%, 1.77
gr/cm3 respectively.
4.2 Strength tests
Purpose of these tests is determination of strength properties
of soil in laboratory conditions. These tests included:- Direct shear test (ASTM D3080) - the results have showedin table I.
- One-dimensional consolidation test (ASTM D2435) – The
tests results have collected in table II.
- Test for measurement of collapse potential of soil (ASTM
D5333) – results showed that the soil of zone havemoderately severe to severe collapse potential that forremoval of this problem, the improvement operation must be
done.
Borehole
NO.
TPF2
TPF5
TPF8
TPF15

Fig. 1- Effect of additives on plasticity of investigated soil

Table I- The results of direct shear tests
Depth ofsample(m)
2
2.2
2
2

Test
method
slow
slow
slow
slow

c (kg/cm2)

φ (deg)

0.09
0.01
0.06
0.03

25
33
28
29

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Table II- The results of consolidation tests

Borehole
NO.
TPF4
TPF5
TPF7
TPF8
TPF13
TPF26

Depth ofsample(m)
2
2.2
2
2
2
2

c (kg/cm2)

φ (deg)

0.13
0.12
0.13
0.10
0.10
0.16

0.005
0.004
0.006
0.006
0.010
0.005

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P

Fig. 2- Effect of cement and bentonite mixture on plasticity ofinvestigated soil

5. Improvement operation
With respect to soil properties of researched zone, thenecessity of soil improvement was quite evident. In continueof this part, admixtures and methods will introduce that havebeen used for modification and improvement of soil
properties.
5.1 Additives
For examining the additives effect on physical and strength
properties of soil, various content of lime, cement, bentonite
and mixture of cement and bentonite and lime and bentonite
was used. Table III shows several mixtures of additives.
5.2 Improvement tests- Atterberg limits test: this test was done for all mixtures ofsoil and additives. Fig. 1 to 3 show additives effect on soil
plasticity.

Fig. 3- Effect of lime and bentonite mixture on plasticity ofinvestigated soil
Results show that cement or lime adding separately doesn't
increase the soil plasticity but increase of bentonite
exclusively, increase the soil plasticity remarkably. Fig. 2 and3 illustrate the effect of additives mixtures on the soil
plasticity. It is evident that the cement and bentonite mixture386

or lime and bentonite mixture increases the plasticity of soil
intensively. Also effect of cement is more than lime.
- Direct shear test: for investigation about the effect ofadditives on strength properties, direct shear test was done on
the new mixture samples. The results have been showed in
fig. 4-7. It is obvious clearly from them that adding thecement have very notable effect on angle of internal frictionof soil. Results show that increase of cement has very large
effect on angle of internal friction than cohesion but it
increases the amount of both.
Adding the lime doesn't have great effect on angle of internal
friction than cohesion but anyway angle of internal friction
has increased but cohesion initially increased and thendecreased. Manner of samples was similar to soft stones andloading on them in these tests doesn't large enough that cause
to compressing the soil structure and active the friction
property between ingredients. Fig. 4-7 show effect of cement
and lime on strength properties of soil.

Fig. 4- effect of cement on cohesion of samples

Table III- additives mixtures
Mix.
NO.

Cementcontent

Bentonite
content

Soil
content

Mix.
NO.

Lime
content

Bentonite
content

Soil
content

1
2
3
4
5

1
1
1
1
1

0
2
5
8
11

99
97
94
91
88

26
27
28
29
30

1
1
1
1
1

0
2
5
8
11

99
97
94
91
88

6
7
8
9
10

3
3
3
3
3

0
2
5
8
11

97
95
92
88
86

31
32
33
34
35

3
3
3
3
3

0
2
5
8
11

97
95
92
88
86

11
12
13
14
15

5
5
5
5
5

0
2
5
8
11

95
93
90
87
84

36
37
38
39
40

5
5
5
5
5

0
2
5
8
11

95
93
90
87
84

16
17
18
19
20

7
7
7
7
7

0
2
5
8
11

93
91
88
85
82

41
42
43
44
45

7
7
7
7
7

0
2
5
8
11

93
91
88
85
82

21
22
23
24
25

9
9
9
9
9

0
2
5
8
11

91
88
86
83
80

46
47
48
49
50

9
9
9
9
9

0
2
5
8
11

91
88
86
83
80

Fig. 5- effect of cement on angle of internal frictionof samples

Fig. 6- effect of lime on cohesion of samples

387

Fig. 7- effect of lime on angle of internal friction of samples

Fig. 10- effect of bentonite and 5% lime on cohesion ofsamplesWith regard of investigations can say that in the samecontent, cement increase has more effect than lime on shear
parameters.
Cohesion of samples increases With increasing the bentonite
content with constant cement or lime content. Adding thebentonite doesn’t has remarkable effect on angle of internal
friction of samples. Fig. 8-11 show these results. In allconditions the cohesion of improved samples has increased
rather than natural soil samples.

Fig. 11- effect of bentonite and 5% lime on angle of internal
friction of samples- Unconfined compressive strength test: The results of thistest have pictured in Fig. 12-14. As a result of adding
chemical additives to soil, several reactions is produced. oneof them that has principal effect on strngth and durability ofsoil, is hydratasion.
The most important properties from soil stabilization withcement additives, is increase of strength with time. In general,
this resistance expresses with parameters such as one
dimensional strength.
It is evident from figures that effect of cement is more thanlime on ultimate compressive strength.

Fig. 8- effect of bentonite and 5% cement on cohesion ofsamplesFig. 9- effect of bentonite and 5% cement on angle of internal
friction of samples
388

Fig. 12- effect of additives on unconfined compressive
strength of samples

Fig. 15- effect of cement on collapse index

Fig. 16- effect of lime on collapse index

Fig. 13- effect of mixture of cement and bentonite on
unconfined compressive strength of samples

The results demonstrate that additives decrease the collapse
potential of soil so that with adding 5% of each additive, soil
loses its collapse potential. Indeed, if the time of production
increases, the effect of additives is more.
6. Conclusion
Regard of this study and obtained results from improvement
operation, these cases result:Fig. 14- effect of mixture of lime and bentonite on unconfinedcompressive strength of samples- Test for measurement of collapse potential of improved
soil: this test was done on samples that were mixed withadditives. Fig. 15 and 16 show the effect of additives indifferent production time on collapse index of improved soil.

1- Grain size distribution tests were done on loess soil ofstudied zone. It is evident from results of tests that soil ofzone is in silty, clayey and sandy loesses limits. Results
emphasize that the more of loess soils are silty.
2- Collapse potential tests were carried out on soil of zone andresults show that the soil was collapsible and its degree wasmoderately severe to severe.
3- Results of atterberg tests on improved samples show that ingeneral existence of bentonite in soil increases its plasticity
extremely so that the plasticity index has increased notably.
Mixtures of cement and bentonite and lime and bentonite also
have caused to increase the plasticity property of soil. But
plasticity index has been decreased by increasing the cement
or lime content in constant content of bentonite because
calcium ion replace with ions on clay mineral surface.

389

Plasticity index on average is 9% in improved soils of Atrakzone.
4- Existence of bentonite, cement and lime and their mixtures
have been caused to changes in cohesion and internal friction
angle resulted from direct shear tests. By this way thatcohesion of soil has been decreased and angle of internal
friction has been decreased generally with increase thecement content or lime content in constant content ofbentonite. But cohesion of soil increased and angle of internal
friction decreased with increase bentonite content in constantcontent of cement or lime. The lowest content of cohesion
was 0.01 kg/cm2 and the most content was 0.09 kg/cm2 in soilof zone before improvement that obtained from direct shear
tests. After adding the mixtures, cohesion content wasincreased to 0.84 kg/cm2. The results of improvement on
loess soils of Atrak zone show that existence of bentonite
together with cement or lime has remarkable effect oncohesion increase.
5- Unconfined compressive strength increases withincreasing the bentonite content in constant content of cement
or lime. Also, it has ascendant procedure with increase
cement or lime content in constant content of bentonite.
6- Adding cement or lime to soil generally causes to removalof collapse potential of the soil. It is resulted that adding about
3% cement or 3% lime is sufficient for removal of thisproperty. Of course the more operation time, the moredecrease the collapse potential.
7- Adding lime and bentonite or cement and bentonite to loesssoil causes to its stabilization that it happens as a result ofchemical reactions between lime and clay or cement or clay.
7. REFERENCES[1] Bell FG, "Engineering Properties of Soil and Rocks," 3rded. Oxford, U. K. Butter worth-Heinemann, 2002, pp.
345.
[2] Clevenger W A, "Experiences with loess as a foundationmaterial," Transactions American Society for Civil
Engineers," vol.123, 1959, pp. 51-80.
[3] Gibbs H. J, Bara J. P, "Stability problems of collapsingsoil. Proceeding American Society for Civil Engineers,"
Journal of Soil Mechanics Foundation Division, vol. 93,
1967, pp. 572-594.
[4] Lutenegger A. J, Hallberg, G. R. "Stability of loess.
Engineering Geology, "vol. 25, 1988, pp.247-261.
[5] Turan Durgunoglu H, etal. "Soil Improvement With Jet
Grout Columns: A Case Study from the 1999 Kocaeli
Earthquake," 5th international Conference on Histories
in Geotechnical Engineering, New York, NY, April
2004.
[6] Rafie B.M.A , et al. "Evaluation of Soil Collapsibility
Potential: A Case Study of Semnan Railway
Station".EJGE. vol.13, 2008.

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Influences of soil material parameters in compaction simulation withsoil/water/air coupled F.E. code
S. Sakamoto, K. Kawai, V. Phommachanh and A. Iizuka
Kobe University, Japan
ABSTRACT
Compacted soil is widely used for earth structures. However, the mechanism of compaction has not been explained in
the framework of soil mechanics up to today. Therefore, it is difficult to grasp distributions of stress, void ratio, and soil
moisture within compacted earth structure. These parameters are needed to predict the behavior of earth structures
exposed to natural disasters. In this study, the influences of permeability and soil water retention characteristics oncompaction quality were obtained from compaction simulations with soil/water/air coupled F. E. code. Here, we
regarded compaction as compression and expansion phenomenon of unsaturated soil under drained air and undrainedwater conditions and formulated it in an initial-boundary-value problem. Consequently, the dependency of thecompaction curve shape on permeability and soil water retention characteristics of soil could be expressed. Moreover,multi-layered compaction was examined.
Keywords: Compaction, Unsaturated soils, Soil water retention characteristics, Permeability
1. INTRODUCTION
Compaction can be defined as pushing air out of soil mass,decreasing voids and increasing dry density. The
compressibility, shear strength and permeability ofcompacted soil strongly depend on its dry density after
compaction. Therefore, compaction control by dry density is
widely applied for designing and constructing compacted
earth structures (Figure 1 and 2). However, soil material is not
uniform and it is difficult to apply the results oflab-compaction tests to geotechnical engineering sites.
Moreover, the effects of randomicity of soil parameters oncompaction quality cannot be estimated since compaction has
not been explained in the framework of soil mechanics.
Kawai et al. (2012) regarded compaction as the compression
and expansion phenomenon of unsaturated soil under drained
air and undrained water conditions, and formulated it as an ininitial-boundary-value problem [2]. They simulated staticcompaction tests with a soil/water/air coupled F. E. code,

named DACSAR-MP [3] and succeeded in expressing thecompaction curve and shear behavior of compacted soil.
In this study, the influences of permeability and soil water
retention characteristics on compaction quality wereconsidered and static compaction simulations withDACSAR-MP were conducted. Moreover, to grasp the
distribution of state quantities, such as void ratio and soil
moisture, multi-layered compaction was simulated.
2. MATHEMATICAL MODEL USED FOR ANALYSIS
The mathematical models applied to DACSAR-MP are
introduced here.
2.1 CONSTITUTIVE MODEL FOR UNSATURATEDSOIL APPLIED TO THE EFFECTIVE DEGREE OF
SATURATION
Ohno et al. (2007) applied suction stress as a part of theeffective stress for unsaturated soil and expressed changes in
stiffness of unsaturated soil with an effective degree ofsaturation as the following equations [4]. The effective stressFig.1 Concept of D-value

Fig.2 Procter’s Concept [1]391is expressed as:=σ ′ σ net + ps 1
where σnet

In DACSAR-MP, the following equations are used for thegoverning equations after the work of Borja (2004) [5].

(1)=σ − pa 1, ps =Se s

S r − S rcs=pa − p w , S e =1 − S rc

(3)Here, σ ′ is the effective stress tensor for unsaturated soil,
σ net is the net stress tensor, 1 is the second rank unit tensor,
σ is the total stress tensor, s is suction, ps is suction stress,
pa is pore-air pressure, pw is pore-water pressure, S r is thedegree of saturation, Se is the effective degree of saturation,and S rc is the degree of saturation at s → ∞ . The yieldfunction is expressed as follows with an EC model which
does not have a singular point on the yield surface.
′, ζ , ε vp ) MD lnf ( σ=nEp′
MD  q +− ε vp 0=′
nE  Mp ′ ζ psat(4)1
λ −κ
, p′ = σ ′ : 1,=
ζ exp (1 − Se ) ns ln a  , MD =
3
1+ e
0

q=

Darcy’s law (pore-water) v w =(6)−k w ⋅ gradh
Darcy’s law (pore-air)(7)v a =−k a ⋅ gradh aContinuity equation (pore-water)(8)
0nSr − S r εv + divv w =Continuity equation (pore-air)p
0 (9)(1 − Sr ) εv + nSr − n (1 − Sr ) a − divv a =pa + p0(2)31
s : s , s =σ ′ − p ′1 =A : σ ′ , A =I − 1 ⊗ 1
2
3Here, nE is the shape parameter, ε vp is plastic volumetricstrain, M is q p ′ on the critical state, D is the dilatancy
coefficient, a and ns are parameters to express expansion of
the yield surface with desaturation, and λ and κ are thecompression and expansion indices, respectively．Figure 3
shows the yield surface expressed as equation (4). The
following elasto-plastic constitutive model for unsaturated
soil is obtained from equation (4) and the associated flow
rule.
σ=′ D : ε − C ⋅ Se(5)

Here, D is the elasto-plastic stiffness matrix, ε is the strain
tensor, and C is the coefficient tensor which expresseschanges in stiffness due to desaturation.
2.2 GOVERNING EQUATIONS FOR PORE-WATER
AND PORE-AIR

Here, v w and v a are the flow velocities of pore-water andpore-air, respectively, k w and k a are the coefficients ofpore-water and pore-air permeability, respectively, h is total
water head, ha is air pressure water head, and p0 is gaugepressure. The soil/water/air coupled problem can beformulated with equations (5), (6), (7), (8) and (9).
2.3 SWRCC MODEL CONSIDERING HYSTERESISThe soil-water retention characteristic curve (SWRCC)
expresses the relationship between suction and saturation. It
is known that SWRCC strongly depends on suction history.
Shibata et al. (2010) showed that this hysteresis causes
distribution of dry density and soil moisture withincompacted soil [6]. In DACSAR-MP, the SWRCC model
proposed by Kawai et al. (2007) [7] is used. This model can
express the difference between drying and wetting processes
as shown in Figure 4.
3 SIMULATIONS OF STATIC COMPACTION
The purpose of compaction is to discharge pore air andincrease dry density. Here, compaction was regarded asloading and unloading under undrained air and drained waterconditions on unsaturated specimens. Static compaction tests
on specimens, which had various permeability and water
retention characteristics, were simulated.
3.1 ANALYSIS CONDITIONS
Figure 5 shows the mesh used for this simulation. Here, 1-D
compression was assumed. Deformation of the bottom was
100

80

i ng
dry

S*

(s1, Sr1)

60 rf

ve
cur

40

S

20 rc

0

100

200

300

400

Suction s (kPa)

Fig.3 Yield surface for unsaturated soil

Degree of saturation Sr (%)

ary
und
Bo

Degree of saturation Sr (%)

100

500

600

*

80Srf(Sra)S(s1, Sr1)

60rfc

*
Sra

Bo
un
da
40
ry
we
tti
ng
Src
cu
20
rv
e
0

100

200

300

400

500

Suction s (kPa)(a)Drying process(b)Wetting process
Fig.4 SWRCC model proposed by Kawai et al.
392

600

Logistic curve Eq.11)
Srf - SraSr =+S
1+exp(A+Blns) ra

0.8

Sra=0.15, Srf=1.0,
BD=5.9, BW=4.6

W

4.0
=-2
AW
.6
-30
A D=
0.8
=-2
AW

0.4

.7
34
=AD

0.6

.2
AD=-25.6
A =-16

Degree of saturation

1

HW
STD
LW

0.2

0
0

Fig.5 Simulation area

λFig.6 Loading condition of compaction

Table 1 Material parameters (STD)

κ

M

200

400

600

80

Suction (kPa)

Fig.7 SWRCC used for simulations
Table 2 Simulation cases
CASE
SUPPLEMENT
STD(standard)－
10k w , 10ka
HP (high permeability)

ν

0.107

0.011

1.344

a

ns

nE

′ (kPa)
psat

0.33

30

1.0

1.3

73.5

LP (low permeability)

0.1k w , 0.1ka

k w (m/day)ka (m/day)me0

HW (high water retention)

2.0 AD , 2.0 AW

0.01

1.0

0.8

0.85

LW (low water retention)

0.5 AD , 0.5 AW

fixed, and that of the left and right boundaries was allowed
only vertically. All boundaries were undrained waterboundaries, and only air drainage from the top of thespecimen was allowed. Compaction was expressed byincreasing and decreasing load on top of the specimen during
compaction. Figure 6 shows the loading conditions insimulations. Actual static compaction was conducted bystrain control and unloading time was less than loading time.
A simple form of loading and unloading was applied. Table 1
summarizes the soil material parameters used for the standard
static compaction simulation (STD). Silty soil was assumed.
Figure 7 shows the soil water retention characteristic curves.
The simulation cases are summarized in Table 2. In order toconsider the influence of permeability, high permeability(HP) and low permeability (LP) specimens are provided.
High water retention (HW) and low water retention (LW)
specimens were also provided. Each soil water retention
characteristic curve is included in Figure 7.
To draw the compaction curve, initial water content was set
bitween 10 to 28% while initial void ratio was 0.85. Initialsuction of all nodes was assumed to take a value in relation to
the initial degree of saturation on the primary wetting curve.
Therefore, vertical distribution of the initial water head was
uneven. Rest time prior to compaction was provided as shownin Figure 6 to allow the water head to reach equilibrium.
3.2 BEHAVIOR OF SPECIMEN UNDER STATICCOMPACTION
Fig.8 shows simulation results on STD. All results are fromthe middle of the specimen (element_3). It was found that the
specimens compressed with loading and expanded withunloading. The yield stress of unsaturated soil increased withdesaturation. Therefore, the time taken to reach a greater
compression point increased with decreased water content(Figure 8 (a)). Since simulations were conducted under
undrained conditions, the degree of saturation increased withcompression and decreased with expansion. Larger
compressive specimens showed larger changes in degree ofsaturation (Figure 8 (b)). Changes of saturation in unsaturated
soil brought about changes in suction. During the loading
stage, the degree of saturation increased due to compression
with a concomitant decrease in suction. On the other hand,during the unloading stage, the degree of saturation decreased
due to expansion and suction increased. These relationships,
between saturation and suction, were dominated by SWRCC(Figure 8 (c)). The influence of this relationship on SWRCC,
was that specimens with higher moisture content caused morechanges in suction. However, suction after compaction was
higher than in specimens with lower moisture content. Also, it
was found that suction after compaction became higher than
before compaction (Figure 8 (d)). This behavior expresses theincrease of strength and stiffness due to compaction.
The aim of compaction is to push out pore-air. Figure 8 (e)shows changes in pore-air pressure. The value of 98kPa
indicates standard atmospheric pressure. Air pressure
increased with loading and decreased with unloading. In
specimens with low water content a relatively low degree ofsaturation produced high air permeability. Therefore pore-air
was smoothly pushed out and returned to 98kPa after
compaction. On the other hand, in high water content
specimens, high air pressure appeared in the loading stage
and little air pressure remained after unloading. This was due
to low air permeability. Pore-water pressure showed the same
behavior as air pressure (Figure 8(f)). However, change in393

w=26%

0.7
w=22%

0.6

w=26%

0.9
0.8

0.6
w=14%

0.5
0.4

Air pressure pa(kPa)

50
w=26%

w=14%
w=18%
w=22%
w=26%

600
400
200

Loading (1min) Unloading (1min)

1.6

w=24%

w=14%
w=18%
w=22%
w=26%

600
400
200
0

Loading (1min) Unloading (1min)

Time t (min)(f) Water pressure

2

w=18%

Height h (m)Height h (m)

0.9

1 .0

w=26%

1

w=20%

w=18%
w=24%

w=22%

1

w=26%

1.5
0

10 12 14 16 18 20 22 24 26 28

0.6

water content w (%)

Fig.10 Distributions of void ratio (HP)2Height h (m)

0.9

1.0

1.6

w=24%

w=26%

1

0.65

0.7

0.7

Void ratio e (-)

Fig.11 Distributions of void ratio (LP)
2

w=18%

w=20%

w=22%

S r=

0.8
S r=

S r=

1.7

0
0.6

0.7

Void ratio e (-)

Fig.9 Influences of permeability
HW
STD
LW

0.65

w=20%

Height h (m)

3

w=20%

w=22%

20

800

(e) Air pressure
Fig.8 Simulation results on STD

S r=

0.8

S r=

S r=

Dry density ρd (g/cm )100

Time t (min)2

Dry density ρd (g/cm 3)

w=14%

0.4
0

Loading (1min) Unloading (1min)(d) Suction

HP
STD
LP

0.5

w=18%

(c) Relationship between
suction and saturation

0Time t (min)

1.7

0.6

Suction s (kPa)

800

w=22%

Unloading

0.7(b) Degree ofSaturation

150

100

w=22%

0.8

Time t (min)(a) Void ratio

w=14%
w=18%

w=26%

0.9

Loading (1min) Unloading (1min)

Time t (min)Suction s (kPa)

w=18%

0.7

Loading (1min) Unloading (1min)

0

w=22%

1

Water pressure pw (kPa)

w=18%

Degree of saturation Sr (-)

Degree of saturation Sr (-)

0.8

1

di ng
L oa

Void ratio e (-)

w=14%

w=24%

w=18%

w=26%

1 w=22%

1.5
10 12 14 16 18 20 22 24 26 28
Water Content w (%)

Fig.12 Influences of SWRCC

0
0.6

0.65

0.7

Void ratio e (-)

0.7

Fig.13 Distributions of void ratio (HW)

pore-water pressure was triggered by change in pore-airpressure since pore-water could not be drained while pore-air
could be drained. This difference appeared as change insuction.
3.3 INFLUENCES OF PERMEABILITY
Since compaction is an initial-boundary-value-problem, it
depends considerably on water and air permeability.
Figure 9 shows the relationship between dry density and

0
0.6

0.65

0.7

0.7

Void ratio e (-)

Fig.14 Distributions of void ratio (LW)

water content in three different permeability specimens,
namely STD, HP and LP. The difference in permeability
appears in the value of dry density. It was found that allcompaction curves had the same optimum water content andcorresponded to each other except in the case of 20-24%
water content. In the lower water content zone of optimum
water content, air permeability was high enough for air to bepushed out. On the other hand, in the higher water content

394

zone of the optimum water content, air was hardly drained
from the specimen. Consequently, the influence of
permeability appeared around the optimum water content. We
can say that the compaction curve has the maximum dry
density according to air permeability. The higher the airpermeability is, the greater the maximum dry density.
Figures 10 and 11 show the void ratio distribution in
specimens HP and LP, respectively. In the high permeability
specimen, HP, void ratio gradually decreased toward the air
drainage boundary. On the other hand, in the low
permeability specimen, LP, void ratio was small only aroundthe air drainage boundary.
3.4 INFLUENCES OF SWRCC
Figure 12 shows the compaction curves in various water
retentivity specimens. We can say that HW is the high waterretentive specimen since it has higher air and water entry
values. The difference bitween the compaction curves
depending on water retentivity can be explained according to
the shape of SWRCC during compaction. In the specimen
with higher water retentivity, the gradient of the wetting
curve on the loading stage was smaller and change in degreeof saturation was smaller than in the lower water retentivity
specimen with the under same water content. Consequently,
air entrapment does not occur easily. Therefore, the optimum
water content is relatively high and the compaction curve
approaches the zero air void curve ( S r = 1.0 ) since entrapped
air did not influence compaction even in the high watercontent region. This characteristic appeared as higher dry
density. However, change in degree of saturation in the highwater retentivity specimen was also small in the unloading
stage and it is difficult to apply this explanation to the totalcompaction curve.
Figures 13 and 14 show the void ratio distributions in
specimens HW and LW, respectively. It was found that the
low water-retentive specimen, LW, was more influenced byentrapped air.
4 SIMULATIONS OF MULTI-LAYERED 1-DCOMPACTION WITH
VARIOUSPERMEABILITY AND WATERRETENTIVITY PARAMETERS
Shibata et al. (2010) simulated multi-layered compaction withsoil/water coupled analysis and expressed the appearance of
discontinuous construction gaps [8]. Here, to discuss theinfluences of soil material parameters on the quality ofcompacted earth structures, multi-layered 1-D compaction
was simulated. The influences of permeability and water
retentivity were also considered.
4.1 ANALYSIS CONDITION
Figure 15 shows the mesh used for this simulation.
“Multi-layered” indicates that the upper layer was createdafter the lower layer had been compacted. In this simulation,
3 layers were used. Each layer had the same thickness anddrainage, and loading conditions were the same as in the staticcompaction simulations.

4.2 RESULTS OF MULTI-LAYERED COMPACTIONSIMULATIONS
Figure 16 shows suction distribution on STD over time. The
time here indicates time after third layer compaction wascompleted. Suction distribution was unsteady anddiscontinuity appeared between the second and third layers
right after compaction. For a while, vertical water balanceoccurred and suction distribution approached hydrostaticpressure distribution. Figure 17 shows the distribution of thevoid ratio. It was found that discontinuous gaps between eachlayer appeared. These gaps were triggered by differences insuction between the lower layer and upper layer. The lower
layer had higher compression generated suction than theupper layer.
4.3 INFLUENCES OF PERMEABILITY
Figures 18 and 19 show void ratio distributions within
multi-layered compactions on high permeability soil, HP, andlow permeability soil, LP, respectively. Since more time wasneeded to reach for equilibrium in the low permeability
sample, LP, changes in void ratio distribution over time canbe seen in Figure 19. However, the final discontinuity gaps
were more remarkable in HP than in LP since it was easier forwater balance between the two layers to occur when the upper
layer was lifted on the lower layer. Consequently, restively
uniform distribution of void ratio appeared in multi-layered
compaction of low permeability soil.
4.4 INFLUENCE OF WATER RETENTION
Figures 21 and 22 show void ratio distributions within
multi-layered compactions in high water-retentive soil, HW,
and low water-retentive soil, LW, respectively. In the low
water-retentive soil, it was easier for the distribution of void
ratio to occur, even within one layer due to the distribution ofdegree of saturation generated by the vertical difference of
total water head. Consequently, discontinuous gaps weremore remarkable in LW than in HW.
5 CONCLUSION
Static compaction and multi-layered compaction were
simulated with soil/water/air coupled F.E code,
DACSAR-MP, and the influences of soil material parameterson the compaction curve and compaction quality wereconsidered. As the result, it was found that soil material
parameters influences appear as a quantitative index of thecompaction curve, such as optimum water content andmaximum dry density.
6 REFFERENCE[1] Procter, R. R. : Four articles on the design andconstruction of rolled-earth dams, Eng. News Record,
Vol.111, pp.245-248, 286-289, 348-351, 372-376, 1933.
[2] Kawai, K., Shibata, M., Ohtsuki, T, Phommachanh, V.,
Kanazawa, S. and Iizuka, A.: Simulations of staticcompaction with soil/water/air coupled F. E. analysis, Proc ofEuropean Conf. on Unsaturated Soils in Napoli, 2012 (inPrinting).
[3] Kanazawa, S.: Estimation of Compacted earth structure’s

395behavior with mathematical model for unsaturated soil,
Doctor Thesis, Kobe University, 2010 (in Japanese).
[4] Ohno, S., Kawai, K. and Tachibana, S.: Elasto-plastic
constitutive model for unsaturated soil applied effective
degree of saturation as a parameter expressing stiffness,
Journal of JSCE, Vol.63, No.4, p.p.1132-1141, 2007 (inJapanese).
[5] Borja, R. I.: A mathematical framework for three-phase
deformation and strain localization analyses of partiallysaturated porous media, Computer Methods in Applied
Mechanics and Engineering, Vol.193, pp.5301-5338, 2004.
[6] Shibata, M., Kawai, K., Kanazawa, S. and Iizuka, A.:
Simulation of static compaction with unsaturated soil/water

coupled F. E. analysis, Proc. 5th Int. Conf. on Unsaturated
Soils, pp.965-971, 2010.
[7] Kawai, K., Iizuka, A., Hayakawa, E. and Wang, W.:
Non-uniform settlement of compacted earth structurescaused by the deformation characteristics of unsaturated soil
on wetting, Soils and Foundations, Vol.47, No.2,
pp.195-206, 2007.
[8] Shibata, M., Kawai, K., Ozaki, S., Kanazawa, S.,
Tachibana, S. and Iizuka, A.: Influences of compaction
procedure upon the quality of compacted earth structures,
Journal of Applied Mechanics, JSCE, p.p.363-370, 2010 (inJapanese).
1.8

1.2

0.6

0
90

Height h (m)Height h (m)

1.8

STD
0day
1day
2days
3days

100

110

1.2

1.2

HP
0day
1day
2days
3days

0.65

Void ratio e (-)

0.7

0
0.6

0.75

1.2

1.2

0

0.6

HW
0day
1day
2days
3days

0.65

0.7

Void ratio e (-)

Fig.20 Changes of void ratio (HW)

396

0.65

Void ratio e (-)

0.7

0.75

Fig.19 Changes of void ratio (LP)
1.8

Height h (m)Height h (m)Fig.18 Changes of void ratio (HP)

Fig.15 Mesh for Multilayered simulation

0.75

LP
0day
1day
2days
3days

0.6

1.8

0.6

0.7

Fig.17 Changes of void ratio (STD)
1.8

Height h (m)Height h (m)Fig.16 Changes of suction (STD)

0
0.6

0.65

Void ratio e (-)

1.8

0.6

STD
0day
1day
2days
3days

0.6

0
0.6

120

Suction s (kPa)

1.2

0.75

LW
0day
1day
2days
3days

0.6

0
0.6

0.65

Void ratio e (-)

0.7

Fig.21 Changes of void ratio (LW)

0.75

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Lower Bound Limit Analysis of Strip Footings Resting on Cohesive Soils
S.M. Binesh, A. Gholampour
Civil and Environmental Department, Shiraz University of Technology, Shiraz, Iran

ABSTRACTA Novel lower bound approach is presented for the bearing capacity determination of strip footings resting on
cohesive soils. The approach consists of the combination of lower bound limit theory and a mesh-free technique. In
the presented method there is no need of mesh in the traditional sense and the constraints at the element
discontinuities are omitted. A statically admissible stress field is constructed by the combination of mesh-freetechnique and the nodal integration scheme. The generated filed is guaranteed to be lower bound by controlling thenon-yielding condition at the entire domain. The solutions of the presented approach are compared with some wellknown solutions, to demonstrate the efficiency and accuracy of the method.
Keywords: Lower bound, Mesh-free method, Strip footing

1. INTRODUCTION
Finding accurate prediction for the bearing capacity of strip
footings is one of the fundamental problems in geotechnicalengineering and, several researches are devoted to the subject[1]-[3]. An appealing choice to solve such problems is theapplication of numerical limit analysis (NLA). The use of
NLA in soil mechanics problems returns back to the study of
Lysmer [4] in which lower bound solutions have been foundfor different stability problems in the geotechnicalengineering. Following Lysmer, Bottero et al [5] introduced a
new finite element formulation for the limit analysis of soil
structures. Sloan [6] combined the Bottero’s approach and theactive set algorithm [7], and developed an efficient method
for the NLA of plane problems in soil mechanics. Several
other attempts have been devoted to the application of NLA
in stability problems in geotechnical engineering [8]-[10].
Up to now, mesh-based methods such as finite element orboundary element method are mostly used in the NLA ofstructures. However, these methods suffer from somedeficiencies which are mainly related to mesh definition. An
alternative approach to get rid of such drawbacks is theimplementation of mesh-free methods as discretization tools.
A great deal of research has been devoted to the applicationof mesh-free methods in different fields of science [11]-[12],however, a few are dedicated to the NLA of structures [13][15]. To the Authors knowledge, there is no study on theapplication of mesh-free NLA in soil mechanics problems
and hence, in present paper, a new mesh-free lower boundformulation is proposed for the bearing capacity of strip
foundations resting on cohesive soils. In this regard astatically admissible stress field is approximated by a meshfree method which uses the Shepard’s shape functions [16].
The stabilized nodal integration technique is adopted toestablish a collocation method for equilibrium satisfaction
throughout the problem domain. The soil beneath thefoundation is assumed to be a cohesive material obeying theTresca yield criterion. Based on the derived formulations a
computer code has been developed and the accuracy andefficiency of proposed method is investigated by solving an
example at the end of the paper.
2. SHEPARD'S METHODThe Shepard's method is used here for the construction ofshape functions. Two distinct properties of Shepard's method
make it appropriate to be used in the lower bound limit
analysis (i) the shape functions constructed by Shepard's
method have the Kronecker delta function property, whichallows the simple imposition of boundary conditions.
(ii) Shepard's shape functions satisfy the maximum
principle. According to this principle, the interpolated values
always lie between the maximum and minimum nodal
values used for the interpolation process. A brief descriptionof the method is presented in this section, and for or moredetails the reader is referred to [17].
Consider a function F ( P ) which is defined over P ⊂ R 2 .
Any finite collection of distinct points in R 2 can be
represented by {Pi }i=1 . The value of F at Pi is shown by Fi
N

and, the Euclidean distance between Pi
point

P

in

R

2

is

denoted

2 1/2

and the generic
by

r

(i.e.

ri = (x - x i ) + (y - yi )  ). Now the function U(P) can2

be written as

397

N 
  NU(P) =  ∑ Fi  ∏ rjα   /  ∑ ∏ rjα  i=1  j≠ i    i=1 j≠ i 

j = 1, 2,..., N(1)

where, α is a positive exponent which can affect the shape where A L is the area of Voronoi cell. Imposition ofof interpolated function. Gordon and Wixom [17] suggested divergence theorem to the obtained equation fromα > 1 for smoothness of interpolated function. In present substituting (7) into (6), leads toresearch α is assumed to be 3.
∂σ ij1
σ ij n j d Γ(8)
By the imposition of nodal values at the N nodes {Pi } , =we∂x j A L ΓL
have:where Γ is the boundary of Voronoi cell and n j is thek = 1, 2,..., NU(Pk ) = Fk(2)normal unit vector in the direction of x j . Equilibrium
The system of equations obtained from (2) leads to∫N

U(P) = ∑ Fφi i (P;P1 , P2 ,..., PN )equation can be rewritten for the smoothed stress gradient by
substituting (8) into (5) as

(3)

i=1

1
σ ij n j d Γ + bi =0
A L Γ∫L

Where φi (P;P1 , P2 ,..., PN ) is the shape function, and can be
written as

φi (P;P1 , P2 ,..., PN=
)

N



∏ rα /  ∑ ∏ rα j≠ ij

 i=1 j≠ i

j



(9)

By satisfaction of equation (9) at all pre-defined nodes,
equilibrium condition for the entire problem domain can beachieved.

(4)

3. LOWER BOUND LIMIT ANALYSISThe lower bound theorem states that the collapse load
obtained from any statically admissible stress fieldunderestimates the true collapse load. A stress field isstatically and plastically admissible if equilibrium andboundary conditions are fully satisfied and the yield
condition is not violated anywhere.
3.1 Equilibrium satisfaction
A mesh-free collocation method is used here in conjunction
with a smoothing technique to satisfy the equilibrium
condition.
Assume a mesh-free scheme for a problem domain in which
the interior domain and the boundaries are constructed bynodes. The general form of equilibrium equations in planestrain condition can be written as

∂σ ij
∂x j

+ bi =
0(5)

where σ ij and bi are the stress tensor component and unit

Figure 1-Voronoi cell around node q
3.2 Boundary conditions
The boundary tractions can be imposed to the problem
solution by the same technique used in section 3.1 forequilibrium satisfaction. In this regard, the gradient of stressin j-direction, in each boundary node Voronoi cell, is set tobe zero. According to Eq. (8) we havebody force respectively.
Equation (5) should be satisfied at all points in the problem∂σ n1
domain, hence a Voronoi cell is constructed around each==
σnn jd Γ 0node (figure 1 ) and the gradient of stress is smoothed over∂x j A LB ΓB
the cell as follows∫∂σ ij
∂x j

=
∫∫ Ψ
ΩL

∂σ ij
∂x j

dΩ

∂τ
1
=
=
τ n jd Γ 0(6)∂x j A LB Γ∫B

where σ ij , Ψ and Ω L are the smoothed stress, smoothing
function and cell domain respectively. According to Chen's
approach [18] the smoothing function can be written as 1Ψ = A L
0

x ∈ ALx ∉ AL(10)

where, A LB and Γ B are related to the boundary node
Voronoi cell. Knowing that the proposed shape functions
have the Kronecker delta function property, the stressboundary conditions can be completely satisfied along theedge by imposing just at the boundary nodes.

(7)
3.3 Yield condition
Tresca yield criterion is adopted here for cohesive soils
behavior. In plane strain condition, this criterion can be
written as398

F =(σ 11 − σ 22 ) − ( 2σ 12 ) − ( 2S u )
2

2

2

(11)

where, S u is the undrained shear strength of cohesive soil.
For a statically admissible stress field we have
(12)
F ≤0
at every points in the problem domain. Equation (12) shows
the locus of points located on and inside a circle in an X-Yplane where X= σ 11 − σ 22 and Y = 2σ 12 . This circle canbe approximated by a polygon of P sides (Fig.2). Thus theyield condition imposes linear inequality constrains on the
stresses as follows:
1, 2,..., PA K σ 11 + B K σ 22 + C K σ 12 ≤ D K =
(13)
where

A k = cos(

4.1 Equilibrium satisfaction
The relation between the smoothed stress gradient and thenodal stress values can be obtained by substituting Eq. (15)
into Eq. (8) as follows:∂σ ij ( x )
∂x j

1
=
=
∑
∫ Φ z ( x ) n j σ ij ( x z )d Γ
A
z ∈K
L ΓL

Substitution of Eq. (16) into Eq. (5), leads to the following
matrix form
Aeq σ = B eq
(17)
where

Aeq = [ A1A2 ... AM ]T(18)

B eq = [ B1 B 2 ... B M ](19)σ = [σ 1 σ 2 ... σ M ]

(20)

T

2πk
)
p

T

Bk = -A k2πk)pπ
D = 2Su cos( )p(14)

Ck = 2sin((16)

where, M is the total number of nodes and, in plane strain
condition, the vector of nodal stresses ( σ i ) and body forces( B i ) can be written as:

σ i = [σ 11 (x i ) σ 22 (x i ) σ 12 (x i )]T(21)

B i = [b1i(22)

b 2i ]

T

where, b1i and b 2i are respectively, the unit body forces in
directions 1 and 2 for node i. The configuration of matrixes
A1 to AM depends on the nodes located in the support
domain of node 1 to M. Suppose that nodes r, s and t arelocated in the support domain of node i, then Ai can be
written as
Ai = [0 ... 0 A re

0 ... 0 Ase 0 ... 0 Ate 0 ... 0]

(23)

where

A e
0
A me 2 Ame =  m 1ee  0 Am 2 Am1 
1A me 1 = ∫ Φ m ( x ) n1d ΓAL Γ

Figure 2-Linearized Tresca failure criterion
4. DISCRETE FORMULATION
According to Eq. (3), the stress values at any point x can be
attributed to nodal stress values as

σ ij ( x=)∑ Φ ( x )σ ( x )

z ∈K

z

ij

z

A

(15)

where, σ ij ( x ) is the stress value at spatial coordinate x ,Φ z ( x ) is the shape function defined by (4), σ ij ( x z ) is thenodal stress value at the spatial coordinate x z , and K is agroup of nodes located in the support domain of point x .
The descritized stress field can be imposed into the requiredconditions for statically and plastically admissible stressfield to derive the discrete form of constraints for lower
bound analysis.

e
m2

(24)

(25)

1
= ∫ Φ m ( x ) n 2d ΓAL Γ

4.2 Boundary conditions
The descritized form of Equation (10) can be obtained bysubstituting (15) into (10) as follows

∑A

1

∑A

1

z ∈K Bz ∈K B∫ Φ ( x ) n σ ( x )d Γ =0
z

LB

n

z

∫ Φ ( x ) n τ ( x )d Γ =0
z

LB

j

ΓB

j

(26)

z

ΓB

where, K B is a group of nodes located in the support
domain of a boundary node, A LB is the area of a boundary
Voronoi cell and Γ B is evaluated at the boundary Voronoi399

cell. Equation (26) can be written for all boundary nodes andthe obtained system of equations can be assembled in thematrix form as
Ab 1σ = 0(27)
Imposition of tractions at the boundary nodes, leads toanother system of equations as follows
Ab 2σ = B b 2(28)
where, Ab 1 and Ab 2 are the coefficient matrixes and B b 2 isthe vector of specified values of tractions along theboundary.
The descritized form of constraints for boundary conditions
can be written as
Abo σ = B bo
(29)
where
Abo =
Ab 1 + Ab 2 , B bo =
Bb 2
(30)

NG

Q = h ∑ ωi σ nf ( x G , y G )where, N G is the number of Gauss points along S , ωi is
the weight of Gauss point i and ( x G , y G ) is the coordinateof Gauss point in ( x , y ) space. To determine the value ofσ nf in Eq. (35), a support domain is considered around each
Gauss point (Fig.3 ). By the application of (15), the final
descritized form of (35) can be written as(36)
Q =C T σ
where C is the coefficient vector and σ is the nodal stressvector.

4.3 Non-yielding condition
Since the discretization method (i.e. Shepard's method)
Posses the maximum principle property, the non-yielding
condition can be checked just at the pre-defined nodes.
According to Eq. (13) the required constrains at all nodes,
can be written in the matrix form as follows:
Ayi σ ≤ B yi
(31)
where,

 A1y0
Ayi = 
 0

 0

0

A2y
0
0

B yi = B1y

B 2y

where

 A1
A
Aky =  2
 

AP

B1
B2

BP

0
0 

0
0 
 0 

0 AMy T B My 

C1 
C 2 
 

CP 

,

(32)

By assembling the obtained relations for the constraints andthe objective function, the problem of finding a staticallyand plastically admissible stress field can be written as

Minimize−C T σSubjected to : Atot σ = B tot(37)

Ayi σ ≤ B yi
where

A=
Aeq + Abo
tot

(33)

,

B=
B eq + B bo
tot

(38)

An inbuilt library program, LINPROG, which is available in
MATLAB, is used for solving above problem.

T

1× P

7. NUMERICAL STUDY

5. OBJECTIVE FUNCTION
For bearing capacity problem the objective function can be
written as

Q = h ∫ σ nf dS

Figure 3- Support domain of a Gauss point under theloading area
6. LINEAR PROGRAMMING PROBLEM( B ) = [ D D  D ]
y
k

(35)

i =1

(34)

S

where, Q is the limit load, h is the thickness normal to theplane, σ nf is the normal stress acted over loaded area on theboundary and S is the length over which the normal stress isexerted. By application of the Gauss method, the integrationin Eq. (34) can be written in a summation form as

In this section, a smooth rigid footing, which is resting on acohesive soil, is considered. The undrained shear strength of
the soil is assumed to increase linearly with depth. The exact
solution of this problem is presented by Davis and Booker[19] as follows:q f = F [(2 + p )S u 0 + r B / 4](39)
where q f is the bearing capacity,

S u 0 is the undrained
shear strength at the ground surface, B is the footing widthandρ is the coefficient of variation of undrained shear

strength with depth.

400

F is a non-dimensional factor whichdepends on the footing roughness and

ρB
Su 0

ρB
Su 0

. By assuming= 3 and S u 0 = 1 , for smooth foundation, F = 1.22

and hence, the exact solution for bearing capacity is 7.1858.
To solve the problem by the proposed method, a mesh-freemodel shown in Fig.4 is considered. The Voronoi diagram
and the boundary conditions are also shown in the figure.
The model has 811 nodes which are oriented in fan pattern.
The support domain around each node is defined by anadjustable method to construct the shape functions. To
guarantee that sufficient and suitable nodes are covered by
the support domains, an automatically self-tuned value isdevised in the code to adjust the radiuses of supports. The
value of bearing capacity obtained from proposed method is7.0324 which is about 2% lower than the exact solution
result.

Figure 4- Problem geometry and boundary condition for
811 nodes and their Voronoi diagrams
8. CONCLUSION
A novel lower bound limit analysis formulation has been
proposed for the bearing capacity determination of strip
footings resting on cohesive soils. In the presented approach,there is no need of mesh in the traditional sense and, theshortcomings related to mesh definition are vanished. The
results of numerical study showed very good lower boundpredictions for the bearing capacity of strip footing resting
on cohesive soil.
9. REFERENCE[1] Bolton MD, Lau CK, “Vertical bearing capacity factors
for circular and strip footings on Mohr–Coulomb soil,”
Canadian Geotechnical Journal 1993;30(6):1024–33.
[2] Frydman S, Burd HJ, “Numerical studies ofbearing-capacity factor N γ ,” Journal of Geotechnical andGeoenvironmental
Engineering(ASCE)
1997;123(1):20–29.

[3] Kumar J, “ N γ for rough strip footing using the methodof characterisitics,” Canadian Geotechnical Journal
2003;40(3):669–74.
[4] Lysmer J, “Limit analysis of plane problem in soil
mechanics,” Soil Mechanics and Foundation Division(ASCE) 1970;96(4):1311-33[5] Bottero A, Negre R, Pastor J, Turgeman S, “Finite
element method and limit analysis theory for soil
mechanics problems,” Computer Methods in Applied
Mechanics and Engineering 1980;22:131-49.
[6] Sloan SW, “Lower bound limit analysis using finite
elements and linear programming,” International Journal
for Numerical and Analytical Methods in Geomechanics
1988;12:61–77.
[7] Sloan SW, “A steepest edge active set algorithm forsolvingsparselinearprogramming
problems.,”International Journal for Numerical Methods
in Engineering 1988;26:2671–2685.
[8] Merifield RS, Sloan W, Yu HS, “Rigorous plasticity
solutions for the bearing capacity of two-layered clays,”
Geotechnique 1999;49(4):471-90.
[9] Kim J, Salgado R, Yu HS, “Limit analysis of soil slopes
subjected to pore-water pressures,” Journal of
Geotechnical and Geoenvironmental Engineering, ASCE
1999;125(1):48-58.
[10] Durand AF, Vargas EA, Vaz LE, “Applications of
numerical limit analysis (NLA) to stability problems ofrock and soil masses,” International Journal of RockMechanics & Mining Sciences 2006;43:408-25.
[11] Atluri SN, Zhu T, “A new meshless local
Petrov-Galerkin (MPLG) approach in computational
mechanics,” Computational Mechanics 1998;22:117-27.
[12] Liu GR, Gu YT, “A point interpolation method fortwo-dimensional solids,” International Journal for
Numerical Methods in Engineering 2001;50:937-51.
[13] Chen S, Liu Y, Cen Zh, “Lower-bound limit analysis byusing the EFG method and non-linear programming,”
International Journal for Numerical Methods inEngineering 2008;74:391-415.
[14] Le CV, Gilbert M, Askes H, “Limit analysis of plates
using the EFG method and second-order cone
programming,” International Journal for Numerical
Methods in Engineering 2009;78(13):1532-55.
[15] Le CV, Gilbert M, Askes H, “Limit analysis of plates andslabs using a meshless equilibrium formulation,”International Journal for Numerical Methods inEngineering 2010;83(13):1739-58.
[16] Shepard D, “A two-dimentional interpolation function
for irregularly-spaced data,” Proc. ACM Nat. conf 1968.
p.517-24[17] Gordon W, Wixom J, “Shepard's method of MetricInterpolationtoBivariate
andMultivariate
Interpolation.,”Mathematics
ofcomputation
1978;32(141):253-64.
[18] Chen JS, Wu CT, Yoon S, You Y, “A stabilized
conforming nodal integration for Galerkin Mesh-freemethods,” International Journal for Numerical Methods
in Engineering 2001;50:435–66.

401

[19] Davis, E. H. , and Booker, J. R., “The effect of increasing
strength with depth on the bearing capacity of clays,”
Geotechnique, Vol. 23, pp. 551–63, 1973.

402

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Upper Bound Limit Analysis of strip footings resting on cohesive soils
S.M. Binesh, S. Raei
Civil and Environmental Department, Shiraz University of Technology, Shiraz, Iran

ABSTRACT
A novel approach is presented for the upper bound bearing capacity determination of strip footings resting on cohesive
soils. The Von-Mises yield criterion is assumed for the cohesive soil behaviour. By the application of associated flow
rule and the normality criterion, the relation between stress and plastic strain increment is assessed at the yield
condition. The radial basis point interpolation method is used to express the strain increment field in terms of nodal
velocities. An optimization problem is established and, solved by an iterative scheme. The efficiency of proposed
method is investigated by solving an example at the end of the paper.
Keywords: Upper bound, Mesh-free, Strip footing

1. INTRODUCTION
The accurate prediction of the bearing capacity of strip
footings is one of the fundamental problems in geotechnicalengineering and, a great deal of research has been devoted tothe subject [1-4]. An alternative approach for solving suchproblems is the numerical limit analysis (NLA) which is arobust method in solving stability problems. The use of NLA
in soil mechanics problems appears to have been first
proposed by Lysmer [5] who found lower bound solutions forthe stability problems in geotechnical engineering. Following
Lysmer, Bottero et al [6] introduced a new finite element
formulation for the limit analysis of soil structures. By the
implementation of Bottero’s approach in conjunction withactive set algorithm [7], Sloan [8] developed an efficientmethod for the NLA of plane problems in soil mechanics.
Several other attempts have been devoted to the application of
NLA in stability problems in geotechnical engineering.
Up to now, mesh-based methods such as finite element orboundary element method are mostly used as discretization
tools in the NLA. However, mesh-based methods sufferfrom some deficiencies which are mainly related to meshdefinition. An alternative approach to eliminate thesedrawbacks in the NLA, is the implementation of mesh-freemethods as discretization tools. There has been a great dealof research into the application of mesh-free methods in
different fields of science [9], however, a few are devoted to
the NLA of structures [10]. To the Authors knowledge,
there is no study on the application of mesh-free NLA in soil
mechanics problems and hence, in present paper, a newmesh-free upper bound formulation is proposed for the
bearing capacity of strip foundations resting on cohesive
soils. In this regard, the Von-Mises yield criterion isassumed for the cohesive soil behaviour and, by theapplication of associated flow rule and the normality
criterion, the relation between stress and plastic strain
increment is assessed at the yield condition. The plasticstrain increment field is expressed by the nodal velocities by
the application of a mesh-free technique called radial pointinterpolation method and, the optimization problem isestablished by considering an objective function andconstrained obtained from upper bound limit analysis theory.
An iterative scheme is proposed for the solution of
optimization problem. The efficiency and accuracy of
proposed method is demonstrated at the end of the paper bysolving an example.
2. UPPER BOUND LIMIT ANALYSISThe theory of upper bound limit analysis is the extension ofplastic work principle [11] by Drucker et al. [12]. Based on
this theory, in an admissible kinematic velocity field, the load
obtained from the equality of internal dissipation power andexternal power, is not lower than the true collapse load. In themathematical form we have

∫T .u d Γ + ∫ f .u d Ω ≤ ∫ σ.ε d Ω
Γ

Ω

Ω

(1)

where T is the traction over surface Γ , u is the velocity
field, f is the body force in domain Ω , σ is the stress fieldand ε is the field of plastic strain rate.
3. YIELD CRITERIONThe Von-Mises yield criterion is adopted here for cohesive
soil behavior modeling. The general form of this criterion canbe written as

403

F=
P −1
(σ ) σ Tσ(2)

where1 1 − 2 0


1  1
P
=
−
1 0
(3)

S u2  2
 0
0 3



in which, S u is the undrained shear strength of the soil.
4. STRESS-PLASTIC STRAIN RATE
The plastic strain used in the upper bound theory should obey
the normality rule. Hence, by considering the associated flow
rule, the plastic strain rate can be written as

ε =µ∂F (σ )

(4)

∂σ

where µ is a non-negative plastic proportionality factor.
Substitution of equation (2) into (4) leads to:ε = 2 µ Pσ
(5)
or

1 −1
σ=
P ε
2µ

where R i ( x ) and Pj ( x ) are respectively, the radial basis
and polynomial terms and, ai and b j are their respective
coefficients. The parameters N and M are the number ofnodes in the support domain and the number of usedpolynomial terms respectively. The velocity field can be
descritized into nodal velocities as follows:u ( x ) = Φ ( x )u S

where u S is the vector of nodal velocity values, and Φ ( x )is the vector of shape functions which can be obtained by
imposing nodal values of velocity into Eq. (10) and tracing astraightforward technique described in [13].

Φ ( x ) = φ1 ( x ) φ2 ( x )  φN ( x ) Tε ( x ) = B ( x )u S ∂φ1
 ∂x 1B (x ) =  0

 ∂φ
 1
 ∂x 2T

For a point at yield surface, where F (σ ) = 0 , the plastic
proportionality factor can be defined asµ=1ε P −1ε
2

P −1ε

(9)

ε P −1ε5. MESH_FREE TECHNIQUE
In this paper, there is no need of mesh in the geometry
modeling and a mesh-free method is implemented as thediscretization tool. In this regard a support domain, whichcontains the neighboring nodes, is defined around each point
and the shape functions are constructed via the nodes locatedin the support domain. The enhanced radial basis pointinterpolation method is used here for the shape functions
construction. Based on this method, a continue function such=u (x )

∂φ2
∂x 1

0



∂φN
∂x 1

∂φ1
∂x 2

0

∂φ2
∂x 2



0

∂φ1
∂x 1

∂φ2
∂x 2

∂φ2
∂x 1



∂φN
∂x 2


0 

∂φN 

∂x 2 
∂φN 

∂x 1 

(14)

It is note worthy that, in the presented mesh-free techniquethe integration process is performed by using a backgroundmesh and the Gauss method and, there is no relation between
the background mesh and the nodal arrangement.
6. CONSTRAINTS REQUIRED FOR ADMISSIBLE
VELOCITY FIELD
A velocity field is admissible if it is kinematically andplastically admissible. In kinematic admissible field thecompatibility condition and the kinematic boundary
conditions should be satisfied. Hence, in plane strain
condition with axes x 1 and x 2 , for kinematic admissibility we
have
(15)
at the entire problem domain
ε = Au

u =0
 ∂
 ∂x
AT =  1

 0


M

∑ R ( x )a + ∑ P ( x )b

i
i
=
i 1=
j 1

j

j

(10)

at the closed kinematic boundaries

(16)

Where

as u ( x ) in a special coordinate x can be approximated as
N

0

(8)

Substituting (8) into (6) gives the relation between stress andplastic strain rate as follows

σ=

(13)

where

Introducing (6) into (2), gives
(7)

(12)

Considering the strain rate and velocity relation, we have(6)

 1 −1   1 −1 
1=P ε PP=ε  −1ε P −1ε − 1
F (σ ) 2
4µ
 2µ
  2µ


(11)

0
∂
∂x 2

∂ 
∂x 2 ∂ ∂x 1 

(17)

The compatibility condition (i.e. Eq. (15)) cab be written as
following relation for descritized velocity field404

ε = Bu S

NG

λ
=
min
ρi ε T P −1ε∑
A velocity field is plastically admissible if the strain ratei =1

vectors belong to the set of normals to the yield surface
subjected to: f .u d Ω + T .u d Γ =1
(normality rule) and the external power of the load is positive.
∫∫ΓThe condition of normality of strain rate vectors, forΩ(25)
Von-Mises criterion leads to=εv 0 on Ωεv = 0(19)=ε Bu Son Ωwhere εv is the volumetric strain rate. The constraint for

=
u 0
on Γ
positivity of external power of load, also results in(20)∫Ω f .u d Ω + ∫Γ T .u d Γ ≥ 0
Since in the limit analysis theory the materials are assumed where N G and ρi are the number and weight of Gauss point(18)

rigid-plastic, and such materials are rate-insensitive, it ispossible to specify the magnitude of velocity without loss ofgenerality. Hence we assume that(21)
f .u d Ω + T .u d Γ =1

∫

∫

Ω

respectively.
For a weightless media, the optimization problem
demonstrated in Eq. (25), can be written in terms of nodal
velocities, using Eq. (18), as follows:
NG

λ
=
min
ρi u ST B T P −1Bu S
∑

i=1


T
subjected to: F u S = 1
Dv Bu S = 0ΓEquations (16),(18), (19) and (21) are considered as
constraints for upper bound limit analysis calculations.
7. ESTABLISHMENT OF MATHEMATICALOPTIMIZATION PROBLEM
To establish the optimization problem, it is assumed that aratio of external load causes the failure of material. In otherwords, Eq.(1) can be modified as follows:(22)
λ ( T .u d Γ + f .u d Ω) ≤ σ.ε d Ω

∫

∫

Γ

Ω

∫

Ω

Where

Ω

(23)

Equation (23) in conjunction with the constraints described in
section 6, leads to the following optimization problem

λ min σ .ε d Ω
=
∫

Ω

subjected to: ∫ f .u d Ω + ∫T .u d Γ =1
Ω
Γ


=
εv 0 on Ω


=
on Ω
ε Bu S


=
on Γ
u 0



Dv = [1 1 0](27)

∫ Φ Td Γ

(28)

=
F

where λ is the coefficient of external loads. Considering
Eq(21), we haveλ ≤ ∫ σ.ε d Ω

(26)

T

Γ

8. OPTIMIZATION ALGORITHM
In this paper the algorithm proposed by Zhang et al. [14] isused for the optimization process. In this algorithm the
Lagrangian method is used in conjunction with theseparation of plastic and rigid zones. In the Lagrangian
method the constraints are combined with the objectivefunction and the Lagrangian function can be written asNG

L (u S , q , µ ) = ∑ ρi u ST B T P −1Bu S

(24)

i =1

(29)

+q ( F T u S − 1) + µ ( Dv Bu S ) =0

Substituting Eq. (9) into (24) and using the Gauss method for
integration calculation, converted Eq. (24) into

where q and µ are Lagrange multipliers. In order to solve
the nonlinear relation shown in Eq. (29), an iterativealgorithm is used. In this regard, an iterative parameter
W icp is defined as

W icp = u ST B T P −1Bu S
and hence, Eq. (29) can be written as

405

(30)

NG

L (u S , q , µ ) = ∑ ρii =1u ST B T P −1Bu S
W icp

NG

u S T B T P −1Bu S
=
λ
min
ρ
∑
i

wk
i =1

T
F uS = 1Dv Bu S = 0 on VTT u S B P −1Bu S = 0 on I nd(31)

0
+q ( F T u S − 1) + µ ( Dv Bu S ) =The iterative algorithm is initiated by the assumption of no
zero strain rates (i.e. a starting value such as 1 for W icp atevery Gauss points), and in the successive steps thenon-differentiable regions are found and isolated by thevalues of W icp . The algorithm can be written as follows:
Initial step :It is assumed that W icp = 1 for every Gauss points. Then, Eq.
31 can be written asNG

L (u S , q , µ ) = ∑ ρi (u ST B T P −1Bu S )

(32)

i =1

The Lagrangian of above optimization problem can be
written as
L=(u S , q , µ ) ∑ ρi u S T B T P −1Bu S + q (F T u S − 1)(37)
+ µ ( Dv Lu S ) + α (u S T B T P −1Bu S ) =0

where α is the penalty factor. Again by differentiation of
L with respect to its variable and, equating them to zero, thefollowing system of linear equations is obtained

 2B T P −1B + 2α B T P −1B F (Dv B )T  u s   0    
00  q  = 1FT00   µ   0 
Dv B
+q ( F T u S − 1) + µ ( Dv Bu S ) =0
According to the Lagrangian optimization theory, by theimposition of

∂L
∂L
∂L
=0,
= 0 and= 0 , following
∂u S
∂q
∂µ

system of linear equations can be obtained.

 2B T P −1BT F Dv B(Dv B )T  u S   0 
   
0  q  = 1
0   µ   0 F
00value of λK +1 can be calculated as follows

λk +1 = ∑ ρi

(33)

i =1

T

T

λk − λk +1
≤ η1
λk +1

(34)

(u S )1 B T P −1B (u S )1

(u S )k +1 B T P −1B (u S )k +1

uk +1 − uk≤η2uk

Step K+1 (K=1,2,…)
From step K, the values of W icp are known for all Gausswhere η1 and η 2 are assumed 0.0001 here.

points and, the rigid and plastic zones can be discerned. Usingsuch definition, the problem domain can be divided into twoparts. I d which indicates the differentiable zones (i.e. plastic)

9. NUMERICAL STUDYand

I nd which is related to non-differentiable regions (i.e.

rigid).

(39)

The iterative calculations are continued until the following
criteria are satisfied

value of λ for the first step can be defined as

λ1 = ∑ ρi(38)

Obtaining (u S )K +1 form above system of equations, theNG

Solution of above system of equations, leads to (u S )1 and the

(36)

(40)
(41)

In this section the well-known problem of undrained loadingof smooth strip footing, is solved by the proposed method.
The exact solution of this problem is given by Prandtl [15]and can be written as

qf = N cSu
(42)
T
I =
T
−1
i ∈ I , (u S )k B P B (u S )k =0(43)
N c= 2 + π
 nd
(35)

T
T
−1
where q f and S u are the exact collapse pressure and the Id =i ∈ I , (u S )k B P B (u S )k ≠ 0undrained shear strength (i.e. cohesion), respectively.
Assuming S u = 1 , the value of exact collapse pressure is

{
{

}
}

Now, the optimization problem at the step K+1 can be
expressed as

equal to 5.14. Fig1. shows the geometry domain and theboundary conditions considered in the present problem. Dueto the symmetry, only one half of the geometry is modeled.
Application of mesh-free technique enables us to consider
different patterns for nodes arrangement. Hence, according to
Fig. 2 three patterns of nodes arrangement are considered for406

the geometry modeling. The problem is solved for eachpattern, and the results are shown in Table 1 As it is obvious,
the fan arrangement of the nodes gives the best results, andthe random arrangement deteriorate the accuracy of results.
However, all obtained results are very good upper boundsolutions for the problem.
To verify the convergence of proposed technique, it isnecessary to show that, by increasing the density of nodes, theaccuracy of results gets better. In this regards, as shown in
Fig. 3, three mesh-free models with fan arrangement of nodes
are considered. As shown in Table 2, all evaluated values areupper than exact solution result and, the obtained trend
confirms the increase of accuracy by decrease of nodes
distances.
a)Fig1. A half-space under uniform normal pressure
Table1. Upper bound solutions for different patterns of nodearrangement.

Meshless

Quantity

NLP

Fan
120 Nodes

Qu/su
No.of iterations

8.769737

regular
120 Nodes
irregular
120 Nodes

Qu/su
No.of iterations
Qu/su
No.of iterations

15.02036
10.70027b)

Table2. Results of upper bound solution for fan arrangement
Meshless
Quantity
NLP
Fine
492 Nodes

Qu/su
No.of iterations

5.170638

Medium248 Nodes
Coarse116 Nodes

Qu/su
No.of iterations
Qu/su
No.of iterations

7.000355
8.632962c)Fig2 Three patterns of nodes arrangement.
a) uniform arrangement b) random arrangementc) Fan arrangement407

constraints related to element definition are omitted. The
results of numerical study show that the proposed method
provides very good upper bound solutions for the bearing
capacity problem.
11. REFERENCES

a)

b)

c)
Fig3. Mesh free models with fan arrangement for a) low
density b) medium density c) high density of nodes
10. CONCLUSION
A novel mesh-free upper bound solution has been proposed
for the bearing capacity determination of strip footings
resting on purely cohesive soils. In the proposed method there
is no need of mesh in the traditional sense and hence, the[1] Liu, Y., Zhang, X. & Cen, Z.,. “Numerical determinationof limit loads for three-dimensional structures using
boundary element method”. European Journal of
Mechanics - A/Solids 2004 ,23, 127–138.
[2] Makrodimopoulos, A. & Martin, C. M., “Upper boundlimit analysis using simplex strain elements andsecond-order cone programming”. International Journal
for Numerical and Analytical Methods in Geomechanics
2006b.,31, 835–865.
[3] Capsoni, A.,. “A mixed finite element model for planestrain limit analysis computations”. Communications in
Numerical Methods in Engineering 1999,15, 101–112.
[4] M. Vicente da Silva , A.N. Antão.,. “Upper bound limit
analysis with a parallel mixed finite element
formulation”. International Journal of Solids andStructures 2008 ,45 - 5788–5804.
[5] J. Lysmer, ‘Limit analysis of plane problems in soil
mechanics’, J . Soil Mech. Found. Div., A.S.C.E., 96(SM4), 2008, 131 1-1334.
[6] A. Bottero, R. Negre, J. Pastor and S. Turgeman, ‘Finite
element method and limit analysis theory for soil
mechanics problems’, Computer Methods Appl. Mech.
Eng. 1980, 22, 131-149.
[7] S. W. Sloan, ‘A steepest edge active set algorithm forsolving sparse linear programming problems’. Int. j.
numer. methods eng. To appear. University of Newcastle,
Civil Engineering Report No. 1987, 022.05.
[8] S. W. Sloan, ‘Upper bound limit analysis using finite
elements and linear programming’, Int. j . numer. anal.
Methods 1989,13, 263-282 /[9] Monaghan JJ. , 1988. “An introduction to SPH”.
Computer Physics Communications; 48:89 -96.
[10] Liu, G.R., Gu, Y.T.,. “A point interpolation method fortwo-dimensional solids”. Int. J. Numer. Meth. Engrg,
Vol.50, 2001,PP.937–951.
[11] Chen, W. F.“Limit analysis and soil plasticity”.
Elsevier Science Publishing Company, Amesterdam.
[12] Drucker, D.C., 1953. Limit analysis of two andthree-dimensional soil mechanics problems. J. Mech.
Phys. Solids 1975, 1, 217–226.
[13] Liu G.R.. “A point assembly method for stress analysis
for two dimensional solids”, Int. J. Solids and Structures,Vol. 2002 ,39, PP.261-276.
[14] Zhang, P.X., Lu, M.W., Hwang, K.,. A mathematicalprogramming algorithm for limit analysis. Acta Mech.
Sinica1991, 7, 267–274 (in English).
[15] Prandtl, L.“U¨ ber die Ha¨rte plastischer Ko¨rper.
Nachrichten von der Gesellschaft der Wissenschaften zu
Go¨ttingen, Mathematisch- Physikalische Klasse 1920
,12, 74–85.

408

Evaluating and optimizing different methods of treating and cleaning
contaminated soil
Dr. Mohammad Reza Atrechian1,Seyed Salman Musavi2 Khosro Ebrahimkhani3
1

Civil Engineering Faculty, Zanjan Azad University, Iran.

2,3

Civil Engineering Faculty, Zanjan Azad University, Iran.

ABSTRACT: One of the main environmental pollutions is soil pollution. Soils are considered as nature
purifies. They also provide food and have refining characteristics. One of the main reasons of soil pollution is
oil leakage and its extension which usually happens near oil refineries and transportation pipes which issometimes unavoidable. These pollutions endanger human health and also the environment. About how to
remove these items from contaminated soils, different solutions that are fully functional and relatively new arepresented. In this article we have been trying to present a new optimized method from different methods aroundthe world according to different criteria such as cost, quality, time and technology. Investigations show that
biological and natural methods of cleaning contaminated soil, is to score higher than other methods.

Keywords: contaminated soil, biological, treatment, optimizing.

mentioned reasons, there are a lot of storage tanks
inside and outside cities which are locatedgenerally underground and sometimes near thewater table. The number of these storages can varywith the city population. As the population
increases the number of these storage tanks
increase too. In places with high water tables thepollution danger is much higher if a leakage occurs
and of course the soil treatment is much harder if
this pollution is happened in a residential area
inside a city.

Introduction
Oil products is one the most dangerous pollutions
which can affect the soil. Oil is usually transported
by vehicles, pipelines and stored in storage tanks.
These are the most common ways which can cause
soil pollution. Leakage is one of the common ways
in polluting the soil especially in fine grain soils
such as sand. The increasing population andindustrialization
has
made
oil
pollution
unavoidable. This phenomena can happen in oil
excavation areas, refineries and in some parts ofpipelines. If the pollution penetrates into deeper
depths the cost and treating operations will
respectively increase and become much harder. Oil
pollution can be treated by three methods: physical,
chemical and biological methods. Biologicalmethods commonly transfer the pollution into
nonpoisonous materials by micro-organism
procedures. This seems to be a more effective
method.

In this paper we have tried to evaluate, optimize
and combine some of the soil treating methods on
different types of soil. Some methods forcontrolling the pollution effects is as follows:
1- Improving the soil ventilation by plowing.
2- Adding nutrients to soil, like Nitrogen andphosphorous.
3- Combining soil with the micro-organisms
which degrade oil products.

Iran is between the top 5 oil producing countries so
it has a high potential of being polluted by oil.
Long pipelines and many oil refineries in south of
Iran are a proof for this potential. Beside the409

non organic compounds. It is usually a pre treating
method just to reduce the contamination so that
other methods can complete the treating procedure.
The contaminated soil is excavated andcategorized, then the coarse grains are separated
and afterwards it is washed by water andsurfactants. This procedure turns the soil into twoparts of clean soil and dirty soil. The advantage ofthis method is that it can gather the contaminations
in settled soil, which this action makes the treatingprocess faster. Soil washing reduces the total costof treatment. Soil washing procedure is closely
related to the soil type of the area. If the grains arecoarse like gravel and coarse sand the process ismore effective and faster than when the soil is finegrained.

Methodology
According to the treatment methods of oil pollution
and its products and due to the previous steps andtreatment procedures discussed in this article, somebiological methods are presented which are matchwith the environmental procedure of soil treatment.
The output of the mentioned procedures are
relatively safe for the biological cycle. Fromdifferent biological methods presented by EPA,
five relatively new methods were randomly
selected and three of them including SVE, BV and
AS were combined to make a relatively new
method.
The five methods mentioned are defined asfollows:
12345-

Soil Flushing: This method is a developed method
for treating, cleaning and gathering thecontaminations of the soil in situ. Soil flushing uses
water and gas compounds to make the geochemical
reactions happen faster which cleans the centralized
contaminations in the underground water.

Soil Vapor Extraction.(SVE)Air Sparging (AS)Bio-venting
Soil Washing
Soil Flushing

METHODS

Generally soil flushing has a high effect on
permeable and homogenous soils such as sand andsilty sand. This method can even treat the harmful
products produced by chemical oxidation and eventreat oil products in soil.

SVE: in this method special preparations are done
to extract toxic and harmful chemical elements
which are formed as gases from soil layers higher
than the water table. These gases are produced
inside the soil and treatment is done by excavatingthese gases.

The New Presented Method
According to the mentioned methods, the offered
theoretical procedure in this paper is defined as Bio
Sparging (BS). As mentioned in the BV method, by
injecting Oxygen (O 2 ) into the contaminated soil
the will be treated. In the AS method by sparging
the air into the contaminated region (which the air
contains O 2 and possibly harmful and miserable
compounds) the treatment is done. Due toenvironmental factors like oil pollutions in theregion (liquid and gas form) , soil type according to
its categorization (which has been categorized in
the 2800 seismic manual of Iran) , depth, physicaland mechanical properties of soil ( permeability
ratio, density, wetness, special mass, organic
material percentage` and etc) the mentioned
nutrients (Nitrogen and Phosphorous) are injected
after being combined with fresh air. For injecting
the air into contaminated soil, two types of wells
are needed, vertical and horizontal wells. The
arrangement of these wells is shown in figure 1).
The horizontal wells must be installed in the

AS: This method is used for treating soils underwater table, by sparging air into soil for evacuating
harmful gases. In this method, air sparging causes
faster evaporation of oil products and afterwards by
SVE the vapors are evacuated.SVE and AS
methods are usually used together for cleaning andtreating a soil.
BV: This method is similar to AS, with thisdifference that instead of air, Oxygen is sparged
into the soil. Oxygen is combined with soil
contaminations and is converted into carbon
dioxide and then excavated from soil. The
advantage of this method towards AS is that in thismethod no harmful vapors are produced and 90 %of the gases return to nature.
Soil Washing: This method is a procedure which is
directly related to water and the treating is
performed out of the contaminated area. Thismethod usually treats soils which have organic and410bottom of the contaminated area with defined
distances from each other in parallel lines and mustbe screened. According to this point that the microorganisms of the natural soil can combine with thecontaminations and change them into natural andbiological gases which reduces the harmfulness ofthese gases to less than 5% can be a goodachievement. Then these gases and harmful vapors
are excavated by SVE wells. Finally these vapors
are taken through carbon filters and are recycled sotheir pollution percentage reduces to the minimum.
Afterwards clean air and sometimes non toxic
water are returned to the nature.

Figure 1: Air Sparging Systemprocedures (for probablepollutions) in the area.

It should be considered that this process iscompletely safe and biological. In this method there
are no chemical surfactants used while injecting
nutrients causes the combination between them andthe micro organism which results in the
improvement of microbes’ growth and increases
the aerification of the treated and cleaned soil.

solid

oil

References:[1] William C.Anderson , 1993, Innovative site
remediation technology, soil washing / soil
flushing. One of an Eight-Volume series.
Edited by:, P.E.DEE. Executive Director,
American Academy of Environmental
Engineers.

Results:
1- This method has a safe and biologicalprocedure because the treating andcleaning materials used in it is O 2 andnutrients (phosphorous and nitrogen).
2- This method which is a combination of thebiological methods in the EPA is an
optimized method.
3- Because of the simplicity of installing theequipment, this method can prepare aproper field for the next cleaning[2] U.S. Environmental Protection Agency ,Off-Gas Treatment Technologies for soil
Vapor Extraction systems: state of the practice,march 2006. Prepared by: Office of superfund
Remediation and Technology Innovation,
Office solid waste and Emergency Response,.

411

[3] Office of solid waste and Emergency
Response (5102G), EPA 542-F-01-006. April
2001.A Citizens Guide to soil Vapor
Extraction and Air Sparging. United states
Environmental protection Agency.

[4] BUDIANTA Wawan, OHTA Hideki,
HINODE Hirofumi, SALIM Chris ,AOKI
Masatoshi , IN-SITU SOIL WASHING
USING INTEGRATED SYSTEM OF
REMEDIATION TECHNOLOGIES. 2007

[5] ARMY, US Army Corps of Engineers
Washington, DC 20314-1000 , Engineeringand Design SOIL VAPOR EXTRACTION
AND BIOVENTING. DEPARTMENT OF
THE.

412

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Experimental Study Concerning Impact Characteristics by Collision of Weight
on Sand Cushion over Steel Beam
Tam Sy HO1, Hiroshi MASUYA2 and Naoto TAKASHITA3
Kanazawa University, Japan
ABSTRACT
Sand cushion is often utilized in protection structures against rockfalls as a shock absorber. Impact by rockfall is
considered one of the most important variable loads for protection structures. To clarify the evaluation method and thebuffering effects of impact force, the series of weight impact experiment were conducted. A simple supported double
steel beam was used to reproduce the behavior of structure. Experiments were carried out by the free fall of the weightto the sand tank installed at the center of simply supported double steel H beams. The impact force of the weight andtransmitted force under the sand tank as well as displacements of beam and strains of the beam were measured.
Dynamic characteristics of impact force, transmitted force to the steel beam and behavior of the beam were
investigated. Dynamic interaction between sand and beam, shock absorbing effect of sand and also the transfer ratio of
kinetic energy of rockfall to the structure were shown.
Keywords: Rockfall, Impact, Sand Cushion, Protection Structure
1. INTRODUCTION
Generally rockfall protection structures are classified into a
net, a fence, a shelf, a wall, embankment, a rockshed etc.
(Fig.1 and Fig.2) [1], [2]. Rockshed is one of the safe andimportant protection structures when the target rockfall has
the large energy. In Japan, many prestressed concreterocksheds and reinforced concrete rocksheds have been
constructed. A certain cushion material is generally installedon the roof of the rockshed for the purpose to buffer the
impact force by a rockfall. Sand is mostly used as typicalcushion material in Japan. Sand and a bag in which sand isFig. 2 Embankment for rockfalls

Net

Rockshed

Fig. 1 Rockshed and protection net

Fig. 3 Failed rockshed by large rockfall

filled are also used for a protection shelf or protection
embankment for the same purpose. Risk is rarely remained
sometimes at the protection structure for the rockfall withlarge energy beyond initial estimation (Fig.3) [3], [4].
Research of this shock absorbing material has been done foryears [5]. However, the evaluation method of impact
behavior and the absorbing effects of impact force which istransmitted to the structure through the cushion are notnecessarily clarified enough. It has been clear empirically thatsand cushion has a large shock absorbing effect. Therefore,
the use of sand cushion material for protection structures,such as a rockshed, is considered rational against the impact

413

force by a rockfall etc. and also in future.
When a protection structure is designed according to the ideaof a performance based design, it is necessary to clarify theultimate state of a protection structure in which sand cushionis installed. It is also required to advance suitable use ofcushion material from the point of repair and reinforcement of
the existing structure. In this research, the series of impact
experiments to the sand cushion on H section steel beam wereconducted in order to obtain the fundamental data about the
impact action for designing a structure safely and rationally.
This paper reports the knowledge acquired by investigating
the impact force, the absorbing effect of sand cushions,dynamic interaction between structure and cushion.

2.2 Measurement Items and Measurement Method
Measurement devices involve an accelerometer (Kyowa
Electronic Instruments Co., Ltd., AS-100HA), load cell
(Kyowa Electronic Instruments Co., Ltd., LUK-1TBS), laser
displacement meter (Keyence Corporation, LB300) and strain
gauge (Tokyo Sokki Kenkyujo Co.,Ltd., FLA-10-11-3-LT) as
shown in Fig.6.
Table 1 Properties of used cushions

2. OUTLINE OF EXPERIMENT
2.1 Method of ExperimentFig. 4 shows the free-fall type device for impact experiment
set up at Structure Engineering laboratory of the Kanazawa
University. The sand tank filled up with sand cushion
material was installed in the center of two H beams, whichwere simple supported locating in parallel. The size of a sandtank was 350 mm in width, 350 mm in depth and 500 mm inheight. It was fixed to H beams with the angle steel beams andbolts.

Type

D10
(mm)

D30
(mm)

D60
(mm)

Effective
particle size
D50 (mm)

Uniformity
coefficientCu

Coefficient ofcurvature
Cc

Sand

0.2

0.34

0.61

0.49

3.10

0.95

Gravel

4.0

5.5

6.5

6.1

1.63

1.16

Concretely, the accelerometer was installed at the center ofthe weight to measure its acceleration. The laser displacement
meter was used to measure deflection of steel beams.
Transmitted force of sand tank to H beams was determinedthrough the load cells placed between the tank bottom and thebeams. The strain gauges mentioned above were stuck tomeasure axial direction strain at the top and bottom flange ofthe central section of H beam.

WeightSand tank

Concrete foundation

H beamFig. 5 Particle size accumulation curve
Table 2 List of impact experiments

Supports
CushionFig. 4 Device of impact experiment
The used H beams were H-100x100x6x8 (mm). The span
lengths of beams were 13 m, 1.8 m, 2.8 m and 3.8 m. The usedweight is a steel cylinder with a diameter of 80 mm and amass of 7.233 kg, and the tip form is spherical. Two kinds ofshock absorbing materials accordingly one kind of sand andone kind of gravel were used. The characteristic values ofsand and gravel are shown in Table 1. The used gravel andsand were crushed stone and loose sand with grain size ranges
as shown in Fig. 5. The sand tank was covered by a thicknessof 0.5 m in both cases.
Table 2 shows the list of all experiments carried out. The
falling heights of the weight were seven kinds respectively
0.5 m,0.75 m, 1.0 m, 1.25 m, 1.5 m, 1.75 m and 2.0 m. 3 times
impact experiments were carried out on each condition.
414

Sand

GravelSpan length ofbeam (m)Falling hights of weight (m)1.50.50,0.75,1.00,1.25,1.50,1.75,2.00

2.00.50,0.75,1.00,1.25,1.50,1.75,2.00

3.00.50,0.75,1.00,1.25,1.5,01.75,2.00

4.0

0.50,0.75,1.00,1.25,1.50,1.75,2.00

1.5

0.50,0.75,1.00,1.25,1.50,1.75,2.00

2.00.50,0.75,1.00,1.25,1.50,1.75,2.00

3.00.50,0.75,1.00,1.25,1.50,1.75,2.00

4.0

0.50,0.75,1.00,1.25,1.50,1.75,2.00

200

Acceleration (m/s2)

Fig.7 shows the measurement system of this experiment. The
output obtained from each measuring instrument wasmeasured at intervals of the sampling of 100 µs (sampling
frequency: 10 kHz) and recorded by PC.
Weight0-200

Acceleration

-400
-600
-800
-1000

0

Acceleration meter

0.02

0.04

0.06

0.08

0.1

Time (s)(a) Acceleration
8

Load cell

6
Force (kN)

Sand tankStrain gauge

Force

4
2
0
-2
-4

0.1

0.08

0.06

0.04

0.02

0

Time (s)H beam(b) Transmitted force (under sand tank)

Stain

Laser displacement meter

Fig. 6 Measurement devices and their locations
Personal computer
Wave logger
Keyence NR500

150 10-6

100 10

-6

50 10

-6

0 10

Strain0-50 10-6-100 10

-6

0

0.02

0.04

0.06

0.08

0.1

Time (s)Measurements unit
Keyence NR-HA08(c) Stain at the span center
Displacement (mm)1.5

DC Amplifier
MA-201ST

1

0
-0.5
-1
-1.5

Accelerometer

Load
cell

Straingauge

Laser
displacement
meter

Displacement

0.5

0

0.04

0.02

0.06

0.08

0.1

Time (s)(d) Deflection at the span center
Fig. 8 Time histories of measured data (Sand, span
length1.8m, falling height 2.0m)

Fig. 7 Measurement system

Pd Ps
or
2
2

3. RESULTS OF EXPERIMENT
3.1 Dynamic behavior of impact experimentFig. 8 shows the time history of measured data for the case
1.8m in span length, 2.0 m in falling height for sand cushion.
The acceleration of the falling weight reached the peak at
approximately 0.01 s after collision to the sand cushion andbecame zero at 0.02 s. Transmitted force measured under thesand tank appeared at 0.005 s, reached its peak at 0.016 s andbecame zero at 0.025 s. The damped oscillation of transmitted
force was shown afterward. Strain and displacement appeared
at 0.01 s, reached its peak also at approximately 0.016 s andbecame zero at 0.025 s. The similar damped oscillations wereshown afterward in both time histories.
Generally the dynamic behavior of structure under hard
impact load is complicated. Here, hard impact means that the
415

Pd Ps
or
2
2
d

Deflection curve

L

a

b

a

M
Bending moment diagram

Fig. 9 Deflection and bending moment of the beam
under equivalent static forces Pd and Ps

magnitude of impulse force is large and duration of impact isvery short. Meanwhile, it is also known that the response ofstructure under relatively soft impact load is mostly
quasi-static. Those beams used in this study have large mass
12
Pa
Pt
Ps
Pd

8

Force (kN)4

at the center of span center. Furthermore, the impact load
occurred by collision of the weight to the cushion is relatively
smooth because of shock absorbing effect of the cushion
material. In that case, it can be assumed that the response of
the beam was quasi-static. Fig. 9 shows the deflection curve
and the bending moment diagram of the simple beam under
two static concentrated loads. Equivalent static forces can be
determined according to the deflection and the strain resulted
by bending moment under this assumption．Here, Ps and

Pd are equivalent forces by the strain and the deflection.

0
-4
-8
-12

0

0.02

0.04

0.06

0.08

L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 mλ=1000 kN/m2
λ=100 kN/m2

0.1

Time (s)(a) Sand (L=1.8m,H=2.0m)
12
Pa
Pt
Ps
Pd

8

Force (kN)40-4(a) Falling height and impact force Pa

-8
-12

0

0.02

0.04

0.06

0.08

0.1

Transmitted force Pt (kN)(b) Sand (L=3.8m,H=2.0m)
6
Pa
Pt
Ps
Pd

4

Force (kN)2

L=1.8 m2.0m
L=2.8 m3.0m

8

L=3.8 m
4.0m

6
4
2
0

0
-2

0

50

100
150
Falling height H (cm)

200

250

(b) Falling height and transmitted force Pt

-4

0

0.02

0.04

0.06

0.08

0.1

L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 mTime (s)(c) Gravel (L=1.8m,H=2.0m)
6
Pa
Pt
Ps
Pd

4
2

Force (kN)

L=1.3 m
1.5m

10Time (s)-6Relation between H and Pt

12

0
-2

(c) Falling height and equivalent force Ps

-4
-6

0

0.02

0.04

0.06

0.08

0.1

Time (s)(d) Gravel (L=3.8m,H=2.0m
Fig. 10 Impact force, transmitted force and equivalent
forces by strain and deflection

Fig. 11 Relationship between falling height andvarious maximum force for sandFig. 10 shows time histories of impact force, transmitted force
and equivalent forces by strain and deflection for four cases,
namely span length L=1.8m and 3.8 m for sand cushion andgravel cushion. The falling height H is 2.0 m.in any case. The
416

impact force Pa is smaller than other forces for sand ( L=1.8m, H=2.0 m). The maximum of force is large in the order of
the transmitted force Pt , the strain equivalent force Ps andthe deflection equivalent force Pd in this case. Meanwhile,the impact force Pa is larger than other forces for sand andlongest span length L= 3.8 m. The shape of first wave oftransmitted force Pt is similar to the impact force Pa .
However, the wave shape of Pt afterward, the equivalent
forces Ps and Pd are similar with longer period than the span
length L= 1.8 m and maximum values of those are smaller
than Pa .For the gravel, similar tendencies are observed.

L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 mλ=1000 kN/m2
λ=100 kN/m2(a) Falling height and impact force Pa
L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 mHowever, maximum values of forces are smaller than those insand. It became clear that the gravel has more excellent shock
absorbing ability than sand and the large flexibility of thebeam itself also contributes to the cushioning role for impact.
3.2 Maximum impact force
Fig. 11 shows the relationship between falling height andvarious maximum impact forces for sand. In Fig. 11(a), linens
estimated by the design formula for the impact load due torock fall are shown. The formula was drawn from the elastic
contact theory and widely used in Japan. This design formulais expressed as following equation [1].
(1)
P = 2.108(mg ) 2 3 λ2 5 H 3 5Here, m is the mass of a falling rock (ton), H is the height of arock fall (m), λ is the Lame coefficient of cushion material(kN/m2) and g is the gravity acceleration (m/s2).
There is no significant difference among span length L
concerning the impact force Pa . The upper limit of the impact
force is expressed by equation (1) of λ=1000kN/m2.
Concerning the transmitted force Pt , it is understood that theforce becomes small, so that the span length is large. Mostly
similar tendency is also observed for the strain equivalent
force Ps . The strain equivalent force Ps is little larger thantransmitted force Pt .
Furthermore, Fig. 12 shows the relationship between fallingheight and various maximum impact forces for gravel. Thereis also no significant difference among span length L
concerning the impact force Pa as observed in the case of sand.
The lower limit of the impact force is expressed by equation(1) of λ=100kN/m2. Concerning the transmitted force Pt , it isunderstood that the force becomes small, so that the span
length is large. Similar tendency is also observed for the strain
equivalent force Ps . The strain equivalent force Ps is smaller
than transmitted force Pt . It is contrary to the case of sand.

(b) Falling height and transmitted force Pt

L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 m

3.3 Dynamic multiplication and energy transfer
It is generally required to estimate the impact load rationallyand safely for the practical design of protection structure.
Some experimental results and discussions are shown here
concerning dynamic multiplication and energy transfer fromthe falling weight to the beam.
Fig, 13 shows the results concerning dynamic multiplication
for sand. Fig 13 (a) shows the relationship between the fallingheight H the dynamic multiplication factor. The dynamic
multiplication factor is generally expressed as following
equation.
Rdyn
(2)
DMF =
Rst
Here, Rst is the response of the structure when the maximum(c) Falling height and equivalent force Ps
Fig. 12 Relationship between falling height andvarious maximum forces for gravel

dynamic force acts statically and Rdyn is the dynamic
response of the structure. In this case, strain is used for the
response of structure. It became clear that there is no
particular relationship between the falling height H and the417dynamic multiplication factor DMF and it is mostly constant.

relationship between Ta T and the dynamic multiplication

The dynamic multiplication factor DMF becomes small if the
span length L of the beam becomes large. Because the longer
span beam has the longer first natural period T for mostly

factor DMF .

constant duration of impact force Ta . Fig 13(b) shows theRelationship between Ta T and dynamic multiplication

Dynamic multiplication factor

factor DMF . It is clearly recognized that there is a proportional

L=1.3 m
L=1.8 m
L=2.8 m
L=3.8 m4 CONCLUSION

1.6
Dynamic multiplication factor

Dynamic multiplication factor(a) Falling height and dynamic multiplication factor

1.4

川砂，1.5m
L=1.3 m1.2
1

川砂，2.0m
L=1.8 m0.8

L=2.8 m
川砂，3.0m

0.6
0.4

L=3.8 m
川砂，4.0m

0.2
0

0

0.5

1

1.5

2

Ta/T

(b) Relationship between Ta /T and dynamic
multiplication factor (Ta : duration of impact force, T :the first natural period of beam )Fig 13 Dynamic multiplication factor by impact (sand)Rate of energy transfer (%)

3.5

Fig. 14 shows the relationship between Ta T and rate ofenergy transfer to beam from falling weight for sand. Theenergy transfer is the transferred energy from the potential
energy of the weight mgH to the beam. It has become clearthat the rate of energy transfer becomes small when theratio Ta T becomes large. The rate of energy transfer REtr isimportant to design the member and structure under the
impact. It can be considered that one of importantcharacteristics of shock absorbing cushion is expressed bythis rate of energy transfer.
We are doing further investigation concerning detailed
formulation of this effect and application to design.

In this research, the series of impact experiments to the sandcushion on H section steel beam were conducted in order toobtain the fundamental data about the impact action.
Obtained results in this research are summarized as follows.
1) The dynamic behaviors of H steel beam with cushion underimpact were concretely shown including characteristic ofimpact force. The concept and actual data concerning
equivalent force were introduced and shown.
2) The impact force Pa obtained by the acceleration of theweight is mostly equal depending only falling height H bothin in sand cushion and gravel cushion.
3) The transmitted force at the bottom of cushion Pt becomes
small when the span length L becomes large because of theeffect of dynamic flexibility of the beam. The force ingravel cushion is smaller than that in sand cushion4) The magnitude of equivalent force Ps or Ps also depends
on the span length as like that of the transmitted force Pt .
5) The dynamic multiplication factor DMF has particular
relationship to the natural period of the beam T and no
relation to the falling height H .
6) The rate of energy transfer to the beam from falling weight
REtr was concretely shown. It has been shown that there isrelationship between REtr and the natural period of thebeam T .

3
川砂，1.5m

L=1.3 m2.5
2

川砂，2.0m
L=1.8 m1.5

L=2.8 m
川砂，3.0m

1

5 ACKNOWLEDGMENT

L=3.8 m
川砂，4.0m

0.5
0

0

0.5

1

1.5

2

Ta/T

Fig 14 Relationship between Ta T and rate of energy
transfer to beam from falling weight (sand)

Mr. A. Satoh and Mr. S. Nakamura of Nihon Samicon co. ltd.
contributed the preparation of experiments. Mr. T. Kojima
and S. Eno, students of Kanazawa University, contributed
effectively to pursuit of experiments. Herein, authors express
sincere great gratitude to them.
REFERENCES[1] Japan road association “Handbook of preventatives
against rockfalls”, Japan road association, 2000 (inJapanese).
418[2] Vogel T., Labiouse V. and Masuya H., “Rockfall
Protection as an Integral Task” Structural Engineering
International, Vol. 9, No.3, 2000, pp.304-3124.
[3] Schellenberg, K. ”On the Design of Rockfall Protection
Galleries”, Doctoral theses, Swiss Federal Institute ofTechnology Zürich, No. 17924, 2009.
[4] Delhomme F., Mommesin M., Mougin J.-P., andPerrotin P., “Behavior of a structurally dissipatingrock-shed: experimental analysis and study of punchingeffects”, Int. J. Solids Struct., Vol. 42, No.14, 2005, pp.
4203–4219.
[5] Masuya H., Aburaya Y., Futo S., Sato A. and Nakamura
S., “Experimental study of the weight collision on a sandcushion and its impact action”, Proceedings of 8thInternational Conference on Shock & Impact Loads on
Structures, 2009, pp. 429-434.

419

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Landslides in Tea Plantation Fields in Shizuoka, JapanJun Sugawara
Golder Associates, Australia
ABSTRACT
Shizuoka Prefecture in Japan is famous for the production of quality Japanese green tea. Approximately 45% of Japan’s
tea is produced in Shizuoka. In this region, tea plants are often grown in hilly terrain. Therefore, due to this topographic
setting, as well as other natural characteristics including geotechnical and geological conditions, tea plantation fields
are occasionally subject to landslides. This paper investigates the relationship between the tea plantation fields andlandslide prone areas in Shizuoka Prefecture. In this study, tea plantation fields are described from the engineering
standpoint. Typical mechanisms of landslides that have occurred in the tea plantation fields are also studied. A series ofinvestigations reveal that there are many common points between the tea plantation fields and the landslide prone areasin this region.
Keywords: Landslides, Tea Plantation Fields


1. INTRODUCTION
Drinking green tea has been a part of everyday life forJapanese people for a long time. It is not only part of theJapanese culture, but it also provides health benefits. A
number of researchers have revealed various potential
positive effects of drinking green tea such as anti-cancer,
anti-oxidant, fat burning, prevention of arteriosclerosis,
lowering blood pressure and reducing cholesterol.
Shizuoka Prefecture in Japan is famous for the production ofquality Japanese green tea. Cultivation of tea plants started
this region in 1241 when a monk named Shoichi Kokushi
planted green tea seeds which he brought back from China [1].
Cultivation of tea plants became economically important in
the 18th century and since then this region leads the tea
production in Japan. Today Shizuoka Prefecture produces
40,000 tonnes of tea, which is approximately 45% of Japan’s
overall tea production, refer Figure 1.

Nara, 2,490t
Kyoto,
2,900tMiyazaki,
3,110tOthers,12,770tMie, 7,230t

Shizuoka,
40,000t

Kagoshima,
23,300t

The author noticed that tea plants are often grown in hilly
terrain and therefore tea plantation fields in Shizuoka areoccasionally subject to landslides. In response to this fact, thispaper investigates the relationship between the tea plantation
fields and landslide prone areas in Shizuoka Prefecture fromthe engineering standpoint.
2. TEA PLANTATION AND LANDSLIDESThe northern part of Shizuoka Prefecture is surrounded by
3,000 meter high mountains which make up a mountain range
called the Southern Alps. Many major rivers in Shizuoka
Prefecture originate in this mountain range (e.g. Fuji River,
Abe River and Ooi River) or the Central Alps (Tenryu River)and flow downstream to the Pacific Ocean. Due to steep river
bed slopes (Fig. 2), these rivers discharge significant volumesof sediment into downstream areas following heavy rainfall
events, causing a number of sediment related disasters. At thetime of April 2012, there are 24,466 designated sediment
related disaster areas in Shizuoka Prefecture [4]. In addition,
there are number of geographical areas named after past
sediment related disasters such as Hatanagi, Ooyakuzure andOokuzure in Shizuoka Prefecture.
Major tea plantation areas in Shizuoka Prefecture, shown in
Fig. 3, overlap somewhat with sediment related disaster areas.
It is known that areas of frequent fog and with a significant
temperature difference between day and night produce highquality tea. Hence many tea plantation fields in Shizuoka
Prefecture are located on the hilly terrain at the foothills of themountains. As a result, tea plantation fields in this region areoccasionally subject to landslides.

Fig. 1 Amount of Tea Production in Japan (after [2])420

Elevation (m)
(m)

1000
800
600

Abe RiverFuji River
Ooi River
Tenryu River

400
200

productive agricultural products using hilly slopes is very
limited. Among these products, tea plantations are considered
as one of the most profitable business [7]. In addition, only
relatively small sized machines, or no machines, are requiredin tea plantation fields making it easy to plant tea trees onslopes.

0
0

100

200

300

400

Distance from River Mouth (km)Fig. 2 Relationship between Elevation and Distance from
Mouth of Class A rivers in Shizuoka Prefecture (after [3])Fig. 3 Major Tea Plantation Fields in Shizuoka Prefecture(after [5])3. MECHANISM OF LANDSLIDES IN TEAPLANTATION FIELDS
3.1 Primary Cause
In Japan, causes of landslides are generally separated into theprimary cause and the triggering cause. The primary causes
are factors related to topography, geology, hydrogeology etc.
and the triggering causes include heavy rainfall, earthquake,
earthworks, and others. In this region, it appears that there aremany common points such as topography, geology, soil
chemistry, hydrogeology and so on between the tea plantation
fields and the landslide prone areas. The following sections
will discuss typical primary cause of landslides in tea
plantation areas.
3.1.1 Topography
In Shizuoka Prefecture, tea trees are often planted on sloping
ground (Photo 1). This is because sloping ground can offer
various advantages over flat ground, including their goodnatural drainage system, longer sunshine duration, larger day
and night temperature difference and lower risk of cold air
stagnation. All these advantages allow production of highquality teas [6] and result in hillsides being the preferred
location for tea plantation fields.
Moreover, advantages of planting tea trees on slopes may not
be limited to these natural conditions. For instance, choice of

Photo 1 Photographs of Typical Tea Plantation Fields on
Sloping Ground in Shizuoka Prefecture
According to [2], 7,377 ha (41% of overall tea plantations in
Shizuoka Prefecture) of tea plantations are on ground inclined
at less than 8 degrees, 4,081ha (24%) is on ground sloping
between 8 and 15 degrees, 5,350 ha (32%) is on slopes steeper
than 15 degrees in Shizuoka Prefecture. Since the total area oftea plantations on slopes steeper than 15 degrees is 6,074 hain Japan, almost 70% of tea plantations on relatively steep
slopes are located in Shizuoka Prefecture. This topographic
characteristic is considered to be one of the primary reasonsthat the chance of landslides in tea plantations in Shizuoka
Prefecture is significantly higher than other prefectures inJapan.

421

Photo 2 Crest of Landslide in Tea Plantation Fields

Photo 3 Tea Plantation Field in Colluvium Slopes

3.1.2 Geology and Soil ChemistryHigh permeability (> 1x10-4 cm/sec), high aeration and highwater retention are important soil characteristics for growing
tea plants. The bedrock of tea plantation fields needs to be
deep and the arable soil, free from impermeable layers, mustbe sufficiently thick. It is also important that the soil mass in
the root system does not contain a large portion of fineparticles.
The Makinohara Plateau located approximately in the centreof Shizuoka Prefecture is one of the most famous tea
production areas in Japan.
This area comprises
approximately 20 to 30 m thick Ooi River alluvial fan
sediments containing Quaternary Pleistocene uncemented
gravelly layers and/or mud layers, which is overlying
Miocene to Pliocene sedimentary layers. While soils in thisregion are suitable for growing tea plants due to their weakly
acidic and high natural drainage characteristics, there are anumber of slope instabilities due to their high erosion
susceptibility.
In addition, colluvium slopes on hilly mountains, which areusually considered not suitable for cultivating agricultural
products, are often used as tea plantation fields. This ismainly because colluvium, which tends to lose basic minerals
and become acidic, is suitable for the growth of tea trees [2].
Tea plants are aluminum-philic and acid resistant. They growwell in the soil in which pH ranges between 4.0 and 5.0.
Nitrogen fertilizer is often used in tea plantation fields since
the quality of the tea increases with an increase of nitrogen in
the soil mass. Because of this, tea plantation fields are often
too acidic for other agricultural products. Therefore little
beneficial effect from tree roots for reinforcing slope stability
is expected in an acidic soil environment. This also
contributes to the fact that tea plantation fields are susceptible
to the erosion during heavy rainfall events.

3.1.3 Past Landslide Sites
Japan has a population of 120 million in an area of about
370,000 km2 and has the one of the highest population
densities in the world. Therefore the redevelopment of pastlandslide sites is an important engineering and social topic in
Japan. To date, past landslide sites are often used as parks
after rectification works in Japan (e.g., Jizukiyama memorial
park and Chausuyama park in Nagano Prefecture).
In Shizuoka Prefecture, tea plantation fields are often foundin the past landslide sites. The author considers that planting
tea trees is a good way of utilising past landslide sites. This isbecause: Profitability of tea plantations per unit cultivable area isgenerally higher than other agricultural products.
 Machines required in tea plantations (e.g., green tea cut
machine) are smaller in size, lighter in weight and lessexpensive compared with those required for otheragricultural products. Therefore they are transportable tomost of the tea plantation fields including difficult accessareas on past landslides and hill slopes.
 Soils in the past landslide sites are often acidic due to lossof alkali ion which suite the growth of tea plants.
In the past landslides sites, tea plantations are often located in
the depression zone near the landslide crown or the lobate
near the landslide toe where the landslides debris aredeposited and slopes become relatively gentle.
It should be noted that these past landslide areas may be usedwithout appropriate rectification works and farmers may beusing these areas without knowing about past landslide events.
Landslides in these areas could be reactivated, even withrelatively minor triggering events.

422

30

300

25

250

20
200
15
150
10

100
5

50
0

Ave.
Monthly
Temperature(Degree(Degree C)
C)
Ave.
Monthly
Temperature

Ave.
Ave.Monthly
MonthlyRainfall
Rainfall(mm)
(mm)

350

Ave. Monthly
Rainfall
Ave. Monthly
Temperature

0
1

2

3

4

5

6

7

8

9 10 11 12

MonthFig. 4 Rainfall and Temperature of Makinohara City (after[9])
Photo 4 Tea Plantation Field in Past Landslide Site

423

BV-3

BV-6
BV-3

Depth to groundwater (m)3.2.1 Rainfall Events and Groundwater
The tea plants can grow well in areas with an average yearlytemperature above 13 degree C, snowfall of not more than 1m
and yearly rainfall of more than 1,400 mm [2]. More than
1,000 mm of rainfall is required during the growth period of
the tea plants between April and October [9]. Many areas in
Shizuoka Prefecture satisfy these requirements due to theirwarm and moist environment (Fig. 4) and thus tea is the oneof the most representative agricultural products in this region.
While rainfall is essential to the growth of tea plants, there are
a number of sediment related disasters due to the heavy
rainfall events. As shown in Table 1, 86 slope failuresoccurred due to heavy rainfall events within a relatively shorttime period.
As described previously, the permeability of the soil in the tea
plantation fields needs to be high. Therefore even if thegroundwater level is low during the dry season, the rise ofgroundwater in tea plantation fields could be significant
during or/and after intense rainfall events (Fig. 5).
Under such circumstances, slope failures may occur suddenly
without any warnings. Once the slope starts moving, the rateof movement could be also very rapid.

Table 1 Summary of Sediment Related Disasters in Shizuoka
Prefecture from 1 Jan. to 3 Oct. 2011 (after [10])Type
Triggering Cause
Earthquake
Heavy Rainfall(incl. Typhoon)
Debris Flow013
Landslides
0
7
Rock falls
7
66
Total
7
86

BV-4
BV-1
BV-5

BV-5
BV-2

Rainfall(mm)3.2 Triggering Cause
Triggering causes of landslides are broadly separated into thenatural causes and those related to human activity. The
natural triggering causes include intense and/or long rainfall
events, snowmelts, earthquakes, scouring or erosion of the toeof the slope and so on. Causes related to human activities
include excavation, filling, tunneling, ponding of reservoirs,
and the like.

Day
Month
Year

Fig. 5 Significant Rise of Groundwater Level in Tea
Plantation Fields due to Rainfall Events
3.2.2 Human Activities
Shizuoka Prefecture has the largest tea plantations fields inJapan. 2.6% of lands in Shizuoka Prefecture are used as tea
plantation fields. Of particular note is that about 25% of
Makinohara City’s land is used as tea plantation fields [9].
Therefore construction activities are often required to extend
into tea plantation fields. As a result, the chance of landslides
in tea plantation fields due to these human activities isinevitably high in Shizuoka Prefecture compared with otherprefectures.

Photo 5 Earthworks Extended Close to Tea Plantation Field

Photo 6 Drainage Well in Tea Plantation Field

4. LANDSLIDE MITIGATION

5. CONCLUSION

Shizuoka Prefecture has approximately 41,200 tea farmhouseholds, which is the highest numbers in Japan. Tea
farming is a very important source of income for them.
Therefore tea farmers are often reluctant to be disturbed byconstruction activities such as landslide rectification works,
particularly during the first tea crop of the season from late
April to May. Governmental administration bodies in thisregion are also fully aware of this issue.
One of the key aspects that needs to be taken into accountduring the selection of landslide rectification works is tominimise the damage to tea plantation fields. Cutting theupper portions of the slope or filling near the toe of the slope,which affects existing tea plantations fields, is therefore often
considered an inappropriate technique by landholders andgovernmental administration bodies.
Moreover when groundwater control measures (e.g., gravity
drains or drainage wells) are adopted for landslide
rectification works, it is also important not to drain
groundwater completely from the tea plantations fields so thatsites remain suitable as tea plantation fields after rectification
works.

In this study, the relationship between the tea plantation fields
and landslide prone areas in Shizuoka Prefecture wasinvestigated. A series of investigations reveal that there aremany common points between the tea plantation fields andthe landslide prone areas in this region. From this study, thefollowing conclusions can be drawn: Almost 70% of tea plantation fields on more than 15 degreeslopes are located in Shizuoka Prefecture. This topographic
characteristic is considered to be one of the primaryreasons that the chance of landslides in tea plantations ishigh in Shizuoka Prefecture.
 Tea plants grow well in soil in which pH ranges between
4.0 and 5.0. Due to this reason as well as meteorologicalconditions, tea plants can grow in colluvial soils which
may be susceptible to the erosion during heavy rainfall
events.
 Tea trees are often planted on existing landslide sites.
Hence these areas are generally susceptible to reactivation.
Landslides in these areas could be reactivated even withrelatively minor triggering causes.
 While rainfalls are essential to the growth of tea plants,
there are numerous sediment related disasters due to heavy
rainfall events.
 Construction activities are often required that extend into
tea plantation fields in Shizuoka Prefecture. As a result, thechance of landslides in tea plantation fields due to thesehuman activities is inevitably high in this region

Shizuoka,
41,200

Others,123,700

Kumamoto,
22,4006. ACKNOWLEDGMENT
Valuable comments were made by Aya Green Tea during
writing of this paper and I am most grateful to them for theirsuggestions.
The author wishes to thank Amanda Barret for reviewing thispaper.

Ibaraki,
15,200Kagoshima,
11,000Miyazaki,
12,600Mie, 12,500

Fig. 6 Number of Tea Farm Households in Japan (after [2])424

7. REEFERENCES[1] Shizuoka Prefecture, “Shizuoka Green Tea”, 2010,
http://www.pref.shizuoka.jp/a_foreign/english/tea/
index.html)[2] Rural Cultural Association, “Cha-Daihyakka”, vol. 1,
Rural Cultural Association, 2008, pp. 269 – 302. (inJapanese)[3] Ministry of Land, Infrastructure, Transport and Tourism,“Kadou-Tokusei”, 2007, http://www.mlit.go.jp/river/basic_info/jigyo_keikaku/gaiyou/seibi/ooi_index.html(in Japanese)[4] Shizuoka Prefecture, “Number of designatedSediment-related Disaster Areas – Municipality Basis”,
2012, http://www.pref.shizuoka.jp/kensetsu/ke-350/
boushi/zyoukyou.html (in Japanese)[5] Teikoku-Shoin, “How to Chizucho”, 2012,
http://www.teikokushoin.co.jp/howto/pdf/index 16/
index16/geo_36.pdf (in Japanese)[6] Shinseiki Koubou, 2007, http://www.cha22.jp/
daimon/001.html (in Japanese)[7] Chamber of Tea Associations of Shizuoka Prefecture,“Ocha Q&A”, 2012, http://www.wbs.ne.jp/bt/chacha/
main/qa_saibai.htm#Q54 (in Japanese)[8] Ito En, “Ocha-Hyakka”, 2006, http://ocha.tv/production/cultivation/ (in Japanese)[9] Makinohara City, “Makinohara City Gaikyo”, 2012,
http://www.city.makinohara.shizuoka.jp/ftp/05gt04/1_g
aikyou.pdf (in Japanese)[10] Japan Sabo Association Shizuoka Branch, “SaboDayori”, vol. 175, Japan Sabo Association Shizuoka
Branch, 2011, pp 2. (in Japanese)

425

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

NATURAL BASE ISOLATION SYSTEM FOR EARTHQUAKEPROTECTION
Srijit Bandyopadhyay1, Aniruddha Sengupta2 and G.R. Reddy 3
Graduate Student & 2 Associate Professor, Indian Institute of Technology Kharagpur, India.
Email: sengupta@civil.iitkgp.ernet.in
3
Professor & Head, Structural & Seismic Engineering Section, Homi Bhaba National Institute, Mumbai, India.
1

ABSTRACTThe performance of a well-designed layer of sand, geo-grid, geo-textiles, and composites like layer of sand mixed withshredded tire (rubber) as low cost base isolators is studied in shake table tests in the laboratory. The building foundation
is modeled by a 200 mm by 200 mm and 40 mm thick rigid plexi-glass block. The block is placed in the middle of a 1m
by 1m tank filled with sand. The selected base isolator is placed between the block and the sand foundation.
Accelerometers are placed on top of the footing and foundation sand layer. The displacement of the footing is alsomeasured by LVDT. The whole setup is mounted on the shake table and subjected to sinusoidal motion with varying
amplitude and frequency. The sand is found to be effective only at very high amplitude (> 0.65g) of motion. Among all
the different materials tested, the performance of a composite consisting of 50% shredded rubber tire and sand placed
under the footing is found to be most promising as a low cost effective base isolator.
Keywords: shake table; base isolation; Shredded rubber tire; model test.

1. INTRODUCTION
To find an economical and feasible way of designing new
structures or strengthening existing ones for protection fromthe damages during an earthquake is one of the challenges in
Civil engineering. The conventional approach to seismic
hazard mitigation is to design structures with adequate
strength and ability to deform in a ductile manner. Over the
past two decades, newer concepts of structural vibration
control including seismic isolation, installation of passive andactive/semi-active devices ([1], [2], [3], [4]) have been
growing in acceptance. Traditionally, earthquake-resistantdesign of low- to medium-rise buildings is particularly
important, as their fundamental frequencies of vibration arewithin the range where earthquake-induced force(acceleration) is the highest as found during Mexico City
Earthquake [5]. One possible mean to reduce the degree of
amplification is to make the building more flexible [6]. In a
low-to-medium-rise building, this necessary flexibility can beachieved by the use of base isolation techniques.
The primary mechanism for the reduction of shaking level ina base isolation method is energy dissipation. The concept of
low-cost and effective earthquake protection techniques using
natural material like sand was looked at by Qamaruddin and
Ahmad [7], Qamaruddin et al. [8] and Feng et al. [9]. The useof a synthetic liner consisting of an ultra molecular weight
polyethylene nonwoven geotextile, placed in the foundationof a structure, was also found to be an effective way of
reducing seismic ground motion by Yegian and Kadakal [10],and Yegian and Catan [11]. Soil reinforced with rubber
demonstrates a tremendous increase in energy dissipation
capability [12]. The feasibility of using shredded rubber
mixed with sand as a natural base isolator was investigatedtheoretically by Gray et al. [13] and Tsang [14]. This paper
presents results of experimental investigations into theperformance of a layer of sand, geogrid, geotextile and sandmixed with shredded rubber tire as low cost base isolation
systems.
2. EXPERIMENTAL SETUP
The laboratory model tests are performed on a 1m by 1m
shake table. The table is shaken in a uniaxial horizontal
direction by specifying a sinusoidal motion of given
amplitude and frequency. The details of the shake table andits calibration can be found in Giri and Sengupta [15]. In thelaboratory model tests, the building foundation or footing isassumed to be square and modeled by a 200mm by 200mm
and 40mm thick, rigid plexi-glass block. The surcharge load
(normal load) on the foundation due to the super structure is
imparted by a number of steel plates (weights) bolted on topof the plexi-glass block. In all the cases, the surcharge load is
15kg. A coarse sand paper is glued to the bottom side of theblock to model the roughness of the model footing. In thelaboratory shake table tests, the model footing is placed inside
a 1000mm by 1000mm and 500mm high, open plexi-glass
container or tank. The plexi-glass container is made up of 12mm thick plexi-glass sheets and reinforced with steel angles
at all the corners and edges. About 30mm thick thermocol
sheets are glued to all the sides of the container except one tominimize the reflection of waves at the ends. One side of the426plexi-glass container is kept clear to monitor the behavior of
the model footing during a test. The whole setup is placed on
top of the shake table and securely clamped to it to ensure no
relative movement. The test container is then filled with sandup to 200mm height and compacted to the required density.
Details of shake table arrangement are shown in Figure 1
below.

different frequencies considered are 1.5, 3.5, and 4.5Hz. For
each specified motion, vertical and horizontal acceleration of
the shake table in addition to those on top of the foundation
sand layer, and on top of the model footing are also recorded.
Each of the motions is continued for at least 5 numbers ofcycles to ensure the system had reached a steady state
condition. A typical input base motion is shown in Figure 2.
In all the cases, the results for the first few cycles are only
shown for the clarity of the presentations. In all the cases, theresults for the first few cycles are only shown for the clarity ofthe presentations.
4. EXPERIMENTAL RESULTS
4.1 Model footing resting on top of foundation sand

The first series of tests are performed with the model footing
placed on top of the sand layer at the middle of the container.
The second series of tests are performed with a geogrid and a
geotextile placed between the model footing and the sandfoundation in the test tank. The third series of the tests are
performed with the shredded rubber tire and sand mixture indifferent proportions as base isolator under the model footing.
Before these third series of tests, a 20mm deep squareexcavation in the sand of the same size as that of the modelfooting is constructed. This excavation is then filled with theshredded rubber tire and sand mixture. The model footing isthen placed over the shredded tire and sand mixture. Several
proportions of shredded tire in the shredded rubber tire-sandmixture have been considered. But only the performances ofsand mixed with 50% shredded rubber tire have been reported
here as it yielded the best result.

Transmitted Acceleration (g)Figure 1. Shake table test setup.

The performance of sand as a base isolator has been studied
for the given base motions mentioned earlier. In this case, themodel footing is resting directly on top of 200 mm deep sandlayer within the test tank. The sand used in the study is a local
uniform medium sand (Kansai River sand). It is classified as
poorly graded sand (SP) as per Unified Soil Classification
System. The specific gravity of the sand is 2.7. The maximum
and minimum dry unit weights are 16.6 and 14.1 kN/m3,
respectively. In all the tests, the relative density of the sandfoundation within the test chamber is maintained at 65%. The
shear strength (effective cohesion, c’ and effective friction
angle, φ’) of the sand, as obtained from the laboratory directotests, are given by c’= 0 and φ’ = 36 . Figure 3 shows thetransmitted peak accelerations at the top of the footing restingon sand with respect to the peak acceleration of the basemotion for different amplitude of motions (keeping thefrequency constant at 3.5 Hz).

0.4

Acceleration (g)

0.3
0.2
0.1
0
-0.1
-0.2

1.2
0.9
0.6
0.3
0
0

0.3

0.6

0.9

1.2

1.5

Peak Base acceleration (g)

-0.3
-0.4

1.5

0

0.2

0.4

0.6

0.8

Figure 3. Footing response for different base motions withconstant frequency of 3.5Hz.

1Time (Sec)Figure 2. A typical input base motion.
3. BASE MOTIONSThe shake table along with the experimental setup is shaken
in horizontal direction by a sinusoidal motions of amplitude
0.15g, 0.3g, 0.4g, 0.6g and 0.8g. The frequency of the motion
is varied from test to test to study the effectiveness of theseismic isolators at different frequency of motion. The

The figure shows that at and around 0.6g amplitude of basemotion, the sand beneath the model footing starts to dampen
the base motion. This is accompanied by a sliding movementof the model footing. This back and forth sliding movementof the model footing is around 3mm in case of 1g motion of
the shake table. It is clear from the above figures and tables
that the sand layer in the foundation behaves as an effective
base isolator only at high amplitude (above 0.6g in these427

4.2 Model footing resting on top of Geogrid & Geotextile
In this case a geo-grid/geotextile is placed underneath themodel footing and on top of the sand foundation. A polyester

geogrid (SF20), composed of high molecular weight,
high tenacity (ultimate tensile strength= 40MPa at 4%
strain), multi-filament polyester yarns, and having
19mm openings is utilized here. The geotextile utilized
here is a 200gsm nonwoven geotextile. The ultimate
tensile strength of the material is found to be 2.3MPa.
The following figure shows the geogrid and geotextile
utilized in this study. The size of the both geogrid andgeotextile are 200mm by 200mm, that is, same as thatfor the model footing.

footing. The response of the model footing resting on thegeotextile is marginally better than the footing resting on
geogrid. As may be found from Figure 6 below, the peakacceleration transmitted to the top of the footing are 0.14g and
0.48g for the corresponding base motion of 0.15g and 0.5g,
respectively.
1.5
Transmitted Acceleration on Footing (g)cases) of base motions.

1.2
0.9
0.6
0.3
0

0

0.3
0.6
0.9
1.2
Peak Base Acceleration (g)1.5

Figure 6. Response of model footing resting on geotextile forbase motions with constant amplitude of 3.5Hz.
(a)(b)
Figure 4. (a) Geogrid & (b) Geotextile utilized in the study.

4.3 Model footing resting on top of a mixture of shredded
rubber tire and sandFigures 5 and 6 show the responses of the footing resting on
geogrid and geotextile to various base motions with constantfrequency of 3.5Hz.

Transmitted Acceleration on Footing (g)

1.5

1.2

In this case, a 20 mm thick layer of sand and shredded rubber
tire mixture is placed between the model footing and thefoundation sand layer within the test tank. The shredded
rubber tire is obtained from a local shop. The average lengthof a thread of shredded tire is 10 mm and the average diameteris 1 mm. Figure 7 shows a magnified view of shredded rubber
tire and sand mixture.

0.9

0.6

0.3

0
0

0.3

0.6
0.9
1.2
Peak Base Acceleration (g)1.5

Figure 5. Response of model footing resting on geogrid forbase motions with constant amplitude of 3.5Hz.
The above figure shows that the geogrid chosen is actuallyamplifying the base motion transmitted to the top of the

Figure 7. Magnified view of 50:50 sand and shredded rubber
tire mixture.
Three different proportions (by weight) of shredded rubber
tire – 20%, 30% and 50% in sand have been utilized in thisstudy as potential low cost base isolators under the modelfooting. However, the results of 50% shredded rubber

428

tire-sand mixture have been only presented here since thesecases exhibit the best results under the present scenario. The
direct shear strength of dry sand mixed with 50% shredded
rubber tire is given by cohesion, c’=30kPa and friction angle,oφ’=30 . In these cases, a 20 mm deep square excavation in thesand of the same size as that of the model footing isconstructed before the tests. This excavation is then filledwith the shredded rubber tire and sand mixture in correct
proportion. The model footing is then placed over theshredded tire and sand mixture (Figure 8). As done for theprevious cases, in this case also the whole test setup is shaken
on the shake table for the previously stated sinusoidal
motions.

motion. This indicates that the isolating layer consisting ofsand mixed with 50% shredded rubber tire is quite effective in
dampening the cyclic motions. Figure 11 attests to theeffectiveness of the sand mixed with 50% shredded rubber
tire as a base isolator for cyclic motions.

Transmitted Acceleration (g)Figure 10. Relative displacement of the footing for a basemotion of 0.3g and 3.5Hz frequency.

Figure 8. Model footing resting on 50% shredded rubber andsand mixture.
The comparison of the peak acceleration at the top of themodel footing resting on sand mixed with 50 percentage ofshredded rubber tire and the peak acceleration of the input(measured during the tests) base motions for a base motion of
amplitude 0.3g and frequency 3.5 Hz are shown in Figure 9.
The relative displacement of the model footing for this case isfound to be about 1.8 mm and shown in Figure 10.

1.5
1.2
0.9
0.6
0.3
0
0

0.3

0.6

0.9

1.2

1.5

Peak Base acceleration (g)Figure 11. Comparison of the peak transmitted acceleration
on top of the footing resting on 50:50 sand-shredded rubber
tire mixture.
Table 1. Variation of peak acceleration at the top of footing
and shake table for different frequencies of base motions withconstant amplitude of 0.3g.
Frequency(Hz)Peak Base
Acceleration(g)Peak Transmitted
Acceleration(g)

1.5

0.3

0.2

2.5

0.3

0.25

3.5

0.3

0.23

4.5

0.3

0.22

Figure 9. Footing response to a base motion of amplitude
0.3g and frequency 3.5Hz.
The comparison of the peak transmitted acceleration on top of
the footing is also compared graphically with peak baseacceleration in Figure 11. The test results show that, unlike insand, even at small amplitude of base motion, the response of
the model footing is remarkably less than the original base429

A base isolating system can be effective in two ways- 1) byreducing the input motion that the structure is subjected to,and 2) by shifting the predominent frequency of the structurefrom that of its base motion, so that resonance of frequency
can not be achieved. This paper only addresses the baseisolation by dampening of the input motion. Since all the
input motions of the shake table are sinusoidal with a given
frequency, the shifting of the frequency of the model footing
during a test could not be studied. It is hoped to study thisimportant aspect of the base isolation system in the next phaseof the study.

Table 2. Variation of peak acceleration at the top of footing
and shake table for different amplitude of base motions withconstant frequency of 3.5 Hz.
Frequency(Hz)Peak Base
Acceleration(g)Peak Transmitted
Acceleration(g)

3.5

0.2

0.15

3.5

0.3

0.23

3.5

0.4

0.32

3.5

0.5

0.43

3.5

0.6

0.48

3.5

0.7

0.52

6. ACKNOWLEDGMENT

The response of the model footing resting on 50% shredded
rubber tire and sand mixture to 0.3g base motions at various
frequencies is shown in table 1. The effect of frequency on the
response of the footing is not so clear at least for this case andother cases looked at. Table 2 shows the response of thefooting to base motions of various amplitudes keeping thefrequency constant at 3.5Hz. The effect of the base isolationis very clear from this table even at small amplitude of
motion.
The shake table test conducted with higher than 50%
shredded rubber tire in sand shows instability even at verysmall amplitude of base motion. The model footing with 15kg
surcharge load on top starts to wobble at the very initial stageof this test and the test is discontinued.
5. CONCLUSIONSThe sand in the foundation of the model footing is found to beineffective as a base isolator at low amplitude (<0.6g) of basemotion. However at higher amplitude (> 0.6g) of motion, it isquite effective in reducing the motion transmitted to thefooting. At 0.8g and 1g, the accelerations at the top of themodel footing show a remarkable decrease and this decrease
in response is also accompanied by back and forthdisplacement of the footing over the sand foundation. At 1g of
motion, the amplitude of this displacement is observed to beabout 4mm.
The shake table tests with the model footing resting on a
20mm layer of sand and shredded rubber tire show that the
proportion of shredded rubber tire should be 50% (by weight)
to yield a significant favorable results. When the proportionof the shredded rubber tire is 50%, the response of the modelfooting is found to be significantly less than the motion of thefoundation and shake table. When the percentage of rubber isfurther increased, the model footing is found to wobble(unstable) at 0.3g motion.

The partial funding by Bhabha Atomic Research Centre(BARC), Mumbai during the course of this study in the formof scholarship for the first author during his graduate study at
IIT Kharagpur is hereby acknowledged.
7. REFERENCES[1] Soong TT, “State-of-the-art review, active
structural control in civil engineering,”
Engineering Structures, vol. 10, no. 2, 1988, pp.
74-84.
[2] Jangid RS and Datta TK, “Seismic behavior ofbase-isolated buildings: a state-of-the-art review,"
Proceedings - ICE: Structures & Buildings, vol.
110, no. 2, 1995, pp.186-203.
[3] Nagarajaiah S, “Semi-active control of structures,"
Proceedings of Structures Congress XV, ASCE,
Portland, Oregon, 1997, pp. 1574-1578.
[4] Ehrgott RC and Masri SF, “Structural control
applications of an electro rheological device,"
Proceedings of the International Workshop on
Structural Control, Univ. of Southern California,
Los Angeles, 1994, pp. 115-129.
[5] Kelly J M, “Base isolation: linear theory anddesign," Earthquake Spectra, vol. 6, no. 2, 1990,
pp. 234-244.
[6] Paulay T and Priestley MJN, Seismic design ofreinforced concrete and masonry buildings, Wiley,
New York, New York, 1992.
[7] Qamaruddin M and Ahmad S, “Seismic responseof pure - friction base isolated masonry building
with restricted base – sliding,” J. Engg. Res., vol. 4,
no. 1, 2007, pp. 82-94.
[8] Qamaruddin M, Arya AS and Chandra B,“Seismic response of brick buildings with sliding
substructure,” Journal of structural Engineering,
vol. 122, no. 12, 1986, pp. 558-572.
[9] Feng M Qing, Shinozuka
M and Fuji S,“Friction-controllable sliding isolation system,”
Journal of Engineering Mechanics, vol. 119, no. 9,
1993, pp. 1845-1864.
[10] Yegian M K and Kadakal U, “Foundation isolation
for seismic protection using a smooth synthetic
liner,”
Journalof
GeotechnicalandGeo-environmental Engineering, ASCE, vol. 130,
no. 11, 2004, pp. 1121-1130.

430

[11] Yegian M K and Catan M, “Soil isolation forseismic protection using a smooth synthetic liner,”
Journal of Geotechnical and Geo-environmental
Engineering, ASCE, vol. 130, no. 11, 2004, pp.
1131-1139.
[12] Edil T B and Bosscher PJ, “Engineering propertiesof tire chips and soil mixtures,” Geotechnical
Testing Journal, ASTM, vol. 17, no. 4, 1994, pp.
453-464.
[13] Gray D-H and Ohashi H, “Mechanics of fiber
reinforcement in sand,” J. Geot. Engg., ASCE, vol.
112, no. 8, 1983, pp. 804-820.
[14] Tsang HH, “Seismic isolation by rubber-soil
mixtures for developing countries,” Earthquake
Engineering and Structural Dynamics, vol. 37, no.
2, 2008, pp. 283-303.
[15] Giri D and Sengupta A, “Dynamic behavior ofsmall scale nailed soil slopes,” Geotech. Geol.
Eng., vol. 27, 2009, pp. 687-698.

431

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Comparison of Effect of Percent of Compaction, Sand and Cement on The
Amount and Time of Swelling of Bentonite
Mohammad S. Pakbaz and B. Rafiee
Shahid Chamran Univ. of Ahvaz

ABSTRACTThen present paper shows the result of a research work on the effect of percent of compaction, of percent of sand, of
initial water content and of percent of cement on the magnitude and time rate of swelling of a commercial bentonite.
Compacted samples of soil prepared and tested free swell under surcharge pressure of 7 KPa and end of primary freeswell and secondary free swell were measured in a consolidation loading frame. The result indicated that initial percentof compaction only affected the amount of secondary swelling and rate of swelling. The results also indicated that
initial water content of 6 % wet of optimum affected the amount of primary swelling and time rate of primary andsecondary free swell but the amount final secondary swelling after 4 days of swelling was not affected by initial watercontent. The result also showed that the percent sand decreased the amount of primary swell and the time to end ofprimary swell. The amount of primary swell and time to primary swell decreased with the percent cement.
Keywords: Four or Five Keywords Compaction, sand, swelling, bentonite, cement
1. INTRODUCTION
Problem soils are classified as expansive soils, dispersive
soils, collapsible soils, liquefiable soils and organic soils (1).
Expansive soils are soils that contain enough amount ofmontmorilonite clay mineral that makes them absorb waterand expand. The most reliable method for evaluating theamount of expansion is direct measurement. The US Burue ofsoil conservation evaluated for the first time the problem of
expansion in soils in 1938. From that time civil engineers
acknowledged that the reason for many building damages
occurred were due to soil expansion (2). Attension to
expansive soils and damages due to theses soils increased
during 1950,s. At that time many housing projects were being
constructed and a lot of problems due to soil expansion werereported that caused many damages to the buildings (1).
As it was mentioned soil expansion is due to the presence ofclay mineral montmorillonite. More than 2.4 million sq.km of
the world land are covered by soils that contain this mineral(2). Due to importance of the subject and the presence of
expansive soil in Khuzestan province of Iran and high groundwater table in this area the problem of soil expansion issevere.
2. IDENTIFICATION METHODS
Direct identification of expansion potential means direct
measurement of percent swelling or swelling pressure. There
are mainly three direct method of evaluating the swelling
potential namely (a) free swell (b) swell under surcharge and(c) constant volume swelling pressure. In this research weused method (a) for our evaluation.

3. MATERIALS AND PROCEDURES
Bentonite- the bentonite we used was available commercially,it had bright grey color with liquid limit, plastic limit,
maximum dry density, and optimum water content of 170 %,
77%, 13.4 KN/m3 and 32% respectively.
Cement- the cement used was type II Portland cement
manufactured in Khuzestan cement factory.
Sand-the sand used was wind blown sand available in
Khuzestan and the distribution curve is shown in Fig.1.
In order to prepare soil samples for free swell tests we mixed
bentonite, sand and cement according to Table 1. For each
mixtures maximum dry density, optimum water content,liquid limit and plastic limit were determined. The samples
were all prepared at maximum dry density and optimum
water content. The samples prepared at dry densities lower
than maximum had water content 2% and 6% dry and wet of
optimum water content. After compaction each specimen
were trimmed into consolidation ring and placed in the
consolidation cell. The water was added to the specimens andthe specimens were allowed to free swell under seatingpressure of 7 KPa. The swelling were allowed beyond the endof primary swelling into secondary swelling.
4. TEST RESULTS
Figure 2 showed the way we determined the end of primary
percent swell (€ v100 ), and time to the end of primary swell(t 100 ).
4.1 Effect of Moisture content
Figure 4 showed the effect of initial moisture content ofspecimens on free swell and time of swell. According to thisFigure the amount of swell for specimen with moisture
content 6 % wet of optimum water content was lower thanspecimens with moisture content 2 % wet or dry of optimum
water content.

432

4.2 Effect of Percent of Compaction
According to Figure 3 the effect of initial percent ofcompaction on the free swell is negligible during the primaryswelling. However the free swell during secondary free swelldecreased with decrease in the percent of compaction.
4.3 Effect of percent of sandFigure 5 shows that in general the increase in the percent ofsand caused a decrease in the percent swell and time to
primary swell.

REFERENCES[1] Asghari, F. and Fakher, A “Swelling and dispersion ofsoils from geotechnical engineer point of view" Tehran
University publication, Tehran, 1993.
[2] Chen, F. H. "Foundation, retaining and earth structures,
Mac Graw Hill book Co. New York, 1988.;

4.4 Effect of percent of cement
Figure 6shows that in general the increase in the percent of
cement caused a decrease in the percent swell and time to
primary swell. The time for primary swelling with the percent
cement beyond 4% cement did not change very much. A
major change of time to end of primary swell occurred forpercent cement below 4% as compared with samples without
cement5.

Conclusion

Compacted samples of soil prepared and tested free swellunder surcharge pressure of 7 KPa and end of primary freeswell and secondary free swell were measured in aconsolidation loading frame. The result indicated that initial
percent of compaction only affected the amount of secondary
swelling and rate of swelling. The results also indicated that
initial water content of 6 % wet of optimum affected theamount of primary swelling and time rate of primary andsecondary free swell but the amount final secondary swelling
after 4 days of swelling was not affected by initial watercontent. The result also showed that the percent sanddecreased the amount of primary swell and the time to end ofprimary swell. The amount of primary swell and time to
primary swell decreased with the percent cement.

Fig.2 Method used for defining primary swell and time to
primary swell

1 Hour3530εv100 (%)

εv100
25

‫ﻧﻣﻭﺩﺍﺭ ﺩﺍﻧﻪ ﺑﻧﺩی ﻣﺎﺳﻪ ﺑﺎﺩی‬120‫ﺩﺭﺻﺩ ﺭﺩ ﺷﺩﻩ‬

100
80
60
40

15
10

20
0
0.01

20

0.1

(mm) ‫ﻗﻁﺭ ﺳﻭﺭﺍﺥ ﺍﻟﮏ‬

1

10

5
0

Fig.1 Grain size distribution curve for sand used

logt100

95%

80%

Percentage of compaction(3%<Optimum water content)

433

65%

Influence of different compaction
on swelling of bentonite

1200
t100 (min)1000
t100 (min)1 Hour

1400

1200

800
600

1000
800
600
400
200

400

0

200

6%>Optimom water content

80%

Percentage of compaction(3%<Optimum water content)

65%

Influence of different
on time of swelling of

3%<Optimom water content

Influence of different moisture
on time of swelling of bentonite

Water content (%)
Compaction=95%095%2%>Optimom water content

Fig.4 The effect of percent of initial water content on thetime of primary swellFig.3 The effect of percent of initial compaction on theamount and time of primary swell

1 Day

35
30
εv100 (%)

31

εv100 (%)

30.5
30
29.5
29
28.5
28

25
20
15
10

27.5
27

5

26.5
6%>Optimom water content

2%>Optimom water content

Water content (%)
Compaction=95%

3%<Optimom wat

Influence of differenton swelling of benton

Fig.4 The effect of percent of initial water content on theamount of primary swell

434

0
0

10

20

30

40

50

60

Percentage of sandFig.5 Effect of percent sand on the amount of primary
swell

70

80

1000

t100 (min)t100 (min)

1000

100

10
0

10

20

30

40

50

60

4
6
Percentage of cement

8

10

70

Percentage of sandFig.5 Effect of percent sand on the time of primary swellεv100 (%)

Table1 Index properties of sample tested

35
30
25
20
15
10
5
0
0

2

4

6

Percentage of cement

8

10

12
7 DaysFig.6 Effect of percent cement on the amount of primary
swell

435

ω
(%)

γ d (max)(kN/m3)

93

32

13.4

63

95

30.4

13.65

70

80

29.5

13.72

176

87

89

28.5

13.75

164

106

58

28

13.75

100B 10C

161

101

60

28.0

13.8

100B 10S

138

60

78

28

14.25

100B 20S

136

53

83

27

14.32

100B 30S

119

43

76

24

15.14

100B 40S

107

41

66

23

15.5

100B 50S

108

41

67

22

16.0

100B 60S

112

48

64

21

16.5

100B 70S

106

46

60

21

16.8

‫ﻧﻤﻮﻧﻪ‬

LL(%)

PL(%)

PI(%)

100B

170

77

100B 2C

158

100B 4C

150

100B 6C
100B 8C

R

(opt)

R

12
7 DaysFig.6 Effect of percent cement on the time of primary
swell

100
0

2

R

R

P

P

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Prediction of the Axial Bearing Capacity of Piles by SPT-based and Numerical
Design Methods
Issa Shooshpasha, Ali Hasanzadeh and Abbasali Taghavi
Babol University, IRAN
ABSTRACTThe prediction of the axial capacity of piles has been a challenge since the beginning of the geotechnical engineering
profession. In recent years determining bearing capacity of piles from in-situ testing data as a complement of static anddynamic analysis has been used by geotechnical engineers. The Standard Penetration Test (SPT) is still the mostcommonly used in-situ test and pile capacity determination by SPT is one of the earliest applications of this test. On theother hand, the acceptance of numerical analyses in geotechnical problems is growing and finite element calculations
are more and more used in the design of foundations. In this paper, different approaches for estimating the bearing
capacity of piles from SPT data have been explained and compared with numerical method. Then, comparisons
between numerical and empirical results are presented and discussed.
Keywords: Pile, Axial capacity, Standard Penetration Test, Finite Element Method
1. INTRODUCTION
The prediction of the axial capacity of piles is recognized as
being the area of greatest uncertainty in foundation design.
Several methods and approaches have been developed toovercome the uncertainty in the prediction. The methods
include some simplifying assumptions or empiricalapproaches regarding soil stratigraphy, soil-pile structureinteraction and distribution of soil resistance along the pile.
Therefore, they do not provide truly quantitative values
directly useful in foundation design [1]. Bearing capacity ofpiles can be determined by five approaches: static analysis,
dynamic analysis, dynamic testing, pile load test and in-situ
testing.
Design guidelines based on static analysis often recommend
using the critical depth concept. However, the critical depth
is an idealization that has neither theoretical nor reliable
experimental support, and contradicts physical laws.
Dynamic analysis methods are based on wave mechanics forthe hammer-pile-soil system. The uncertainty in the hammer
impact effect, as well as changes in soil strength from theconditions at the time of pile driving, and also at the time ofloading, causes uncertainties in bearing capacity
determination. Dynamic testing methods can only be used by
an experienced person and the capacity estimation is notavailable until the pile is driven [2].Pile load test is the best
method for pile bearing capacity determination but such tests
are expensive, time-consuming and the costs are often
difficult to justify for ordinary or small projects. In recent
years, the application of in-situ testing techniques has
increased for geotechnical design. This is due to the rapid
development of in-situ testing instruments, an improved
understanding of the behavior of soils.

2. COMMON METHODS of PILECAPACITY DETERMINATION by SPT

BEARING

The Standard Penetration Test, SPT, is still the mostcommonly used in-situ test. Also SPT is the most common
testing methodology adopted in the field to gain idea about
the stratigraphic profile at a site. Due to its simplicity of
execution (apart from the difficulty in repeatability), a fieldengineer finds the method to be one of the most amiable andreliable one.
The introduction in the United States in 1902 of driving a
25-mm diameter open-end pipe into the soil during thewash-boring process marked the beginning of dynamic
testing and sampling of soils. Between the late 1920s andearly 1930s, the test was standardized using a 51-mm O.D.
split-barrel sampler, driven into the soil with a 63.5-kg
weight having a free fall of 760 mm. The blows required todrive the split-barrel sampler a distance of 300 mm, after an
initial penetration of 150 mm, is referred to as the SPT Nvalue. This procedure has been accepted internationally with
only slight modifications. The number of blows for each of
the three 150-mm penetrations must be recorded. Apart fromits main applications in soil characterization, SPT N-value
has also been extensively used for designing structural
foundations and other earth structures, particularly, for the
bearing capacity of piles.
Pile capacity determination by SPT is one of the earliest
applications of this test that includes two main approaches,
direct and indirect methods. Direct methods apply N values
with some modification factors. Indirect SPT methods
employ a friction angle and undrained shear strength values
estimated from measured data based on different theories.
Amongst the two, the direct methods are more acceptedamongst the field engineers for the ease of computations. In
the present study, the following common SPT-based direct436

methods have been employed to predict the pile bearing
capacity (Table I).
Table I: SPT direct methods for prediction of pile bearing capacity in the present study

MethodUnit Base (Q b ) and Unit
Shaft (Q s ) resistancem N b ≤  L  Q b (MPa) = k
R

R

D

Meyerhof(1976)Nb
R

Remarks
N b :average of N between 10D above and 5D below
pile baseN s : average value of N around pile embedment depth.
bored piles: n s =1, k=0.012, m=0.12
driven piles: n s =2, k=0.04, m=0.4
R

R

R

R

R

R

Q s (kPa) = n s N s
R

R

R

R

R

R

Q b (MPa) = n b N b
R

R

R

R

R

R

Q s (kPa) = n s N s
R

R

R

R

R

Q b (MPa)= k b N b
R

R

R

R

R

R

R

R

R

R

N b :average of N between 1D above and 3.75D
below pile base, N b ≤ 50
n b = 0.06 - 0.2
ns= 2 – 4
N s : average value of N around pile embedment
depth
driven piles and bored piles in clay: α =1bored piles in granular soils: α=0.5-0.6
driven piles in sand: k b =0.325
bored piles in sand: k b =0.325
driven piles in clay: k b =0.1
bored piles in clay: k b =0.08
N b : average of N around pile baseN s : average value of N around pile embedment
depth.
N gb : the geometrical average of N values between 8D
above and 4D below pile baseN gs : geometrical average of N values along the pile
R

Bazaraa &Kurkur (1986)

R

R

R

R

R

R

R

Decourt(1995)R

Q s (kPa) = α (2.8 N s +10)
R

R

R

R

R

R

R

R

Shariatmadariet al.(2008)

Q b (MPa)= 0.385 N gb
R

R

R

R

R

Q s (kPa) = 3.65 N gs
R

R

R

3. DETERMINATION of PILE BEARING
CAPACITY by SPT (Case Study)

P

R

R

R

Fig 2. Exploratory boring log

By performing a 15m borehole in a region in Babol (Fig1), north of IRAN, the soil stratigraphy has beenrecognized(Fig 2). The groundwater table is 1.5 m below
the ground surface. The average unit weight of the soil is
1.93gr/cm3 and its average modulus of elasticity is 90
kg/cm2. Fig 3 depicts the variation of SPT-N values withdepth of this site.
P

R

R

R

R

R

R

R

P

P

Fig 1. The region considered in this study437

sediments down to great depths. These tertiary sediments
were stiff plastic clay similar to the so-cal1ed Frankfurt
clay, with a varying degree of overconsolidation. The
groundwater table was about 3.5m below the groundsurface. The considered pile had a diameter of 1.3 m and alength of 9.5m. It was located completely in theoverconsolidated clay. The loading system consisted of
two hydraulic jacks working against a reaction beam. This
reaction beam was supported by 16 anchors. Theseanchors were installed vertically at a depth between 15
and 20 m below the ground surface at a distant of about 4m from the tested pile, in order to minimize the effect ofthe mutual interaction between the tested pile and thereaction system. Then, loading test was carried out. The
loads were applied in increments and maintained constantuntil the settlement rate was negligible.
The mentioned pile load test has been carried out by ElMosallamy in Frankfort clay, was used to verify thepresent numerical method. Fig 4 shows the layout of thepile load test and measurement points by El-Mossallamy.
Fig 5 demonstrates a comparison between the result
obtained by El-Mosallamy and that obtained in thepresent study .The comparison depicts a good agreement
between the results.

Depth (m)SoilClassification(unified)

GraphicLog

LL (%)PI (%)

first
15 cm

second
15 cm

third
15 cm

Total
N

Standard Penetration Test
(No. of Blows)

0
Fill
1

5

3

2

2

20

37

NP
12

6

6

6

6

2

SP-SM

3

4

4
SP-SM

NP
12

CL

4

5
6

SW-SM

NP

7
6
8

4

4

23

CL-ML

9

NP
15

8

7

SM
10

4
SM

NP
12

6

6

4

3

3

11
12

4
SM

NP

7

8

4

16

13

ML

44

14
15

Fig 4. Layout of the pile load test and measurement points
by El-Mossallamy

Fig 3. Variation of SPT-N values with depth

Based on the SPT data above, a hypothetical problem isconsidered wherein the bearing capacity of a bored pile
(L=7m, D= 1m) is determined by the methods mentioned
above.( Table II ).
Table II: The bearing capacity of the bored pile bycommon SPT-based methods (L=7m, D=1m)
MethodMeyerhof

value(kN)916

Bazaraa &Kurkur1102

Decourt1700

Fig 5. Comparison between the results of El-Mosallamy
and the present study (a) load – settlement curve (b) stress– displacement curve

Shariatmadari
3655

4. VALIDATION
In order to validate the program, a pile load test inGermany has been analyzed. The load test investigatedthe load-settlement behavior of a single pile. The upper
4.5m subsoil consist of silt (loam) followed by tertiary
438

(a)
The stress-settlement behavior of the pile, analyzed by the
program is presented in Fig 7.
Fig 7. Stess-settlement behavior of the considered pile(b)
5. DETERMINATION of PILE BEARING
CAPACITY by PLAXIS 3D FOUNDATIONIn this part, the bearing capacity of the pile mentioned
above is determined by the program "plaxis 3d
foundation". "Plaxis 3d foundation" is a finite element
method-based three-dimensional program speciallydeveloped for the analysis of foundation structures,
including off-shore foundations. This program combines
simple graphical input procedures, which allow the user toautomatically generate complex finite element models,with advanced output facilities and robust calculation
procedures. In order to analyze the behavior of the single
pile, at first a model is made. A working area 50m x 50m
has been used. The pile is modelled as a solid pile using
volume elements in the center of the mesh. Interfaces aremodelled along the pile. The generated mesh of thesystem (for pile: L=7m, D= 1m) by this program is asfollows (Fig 6).

The bearing capacity of the considered pile, obtained bythis program, is 1200 kN. Fig 8 shows a comparison
between different methods and the present study.
Fig 8. Comparison between the results obtained bydifferent methods and the present study FEM

Fig 6. The typical mesh used in the study
6. CONCLUSION
Determining the bearing capacity of piles is an interesting
subject in geotechnical engineering. The complex natureof the embedment ground of piles and lack of suitable
analytical models for predicting the pile bearing capacity
are the main reasons for the geotechnical engineer's
tendency to peruse further research on this subject. Direct
bearing capacity predicting methods for piles aredeveloped based on in-situ testing data, specially SPT,
439

having applications that have shown an increase in recent
years.
In this study, by analyzing the pile by FEM andcomparing it with common methods, tried to find areasonable prediction for its bearing capacity. The results
indicate that, in this site, the bearing capacity predicted by
Bazaraa & Kurkur method is very close to the one
obtained by the finite element method (FEM). Also
Meyerhof method underestimates and Shariatmadarimethod overestimates the bearing capacity of the single
pile. Therefore, 1200 kN can be a good prediction for the
bearing capacity of the bored pile considered (L=7m,
D=1m) in this study.
7. REFRENCES[1] Eslami A & Fellenius, BH, “Pile capacity by direct
CPT and CPTu methods applied to 102 case histories,”
Canadian Geotechnical Journal, vol. 34, 1997, pp.
886-904.
[2] Rausche F, Goble G & Likins G, “Dynamic
determination of pile capacity,” J of GeotechnicalEngineering, vol. 111, 1985, pp. 367-383[3] Meyerhof GG, “Bearing capacity and settlement of pilefoundations,” J of the Geotechnical Engineering Division,
ASCE, vol. 102, Iss. GT3, pp.196–228.

[4] Bazaraa AR & Kurkur MM, “N-values used to predict
settlements of piles in Egypt,” Proceedings of In Situ ’86,
New York, 1986, pp. 462-474[5] Decourt
L,“Predictionofload-settlement
relationships for foundations on the basis of the SPTT,” Ciclo de Conferencias Internationale, Leonardo
Zeevaert, UNAM, Mexico, 1995, pp. 85-104.
[6] Shariatmadari N, Eslami A & Karimpour-fard M,“Bearing capacity of driven piles in sands from SPT–
applied to 60 case histories ,” Iranian Journal of
Science & Technology, 2008, vol .32, pp.125-140[7] El-Mossallamy Y “Load-settlement behavior of large
diameter bored piles in over-consolidated clay,”
Proceeding of the 7th. International Symposium
Models in Geotechnical Engineering, Graz, Austria,
1999, pp.443-450

440

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Application of Recently Developed Mechanistic-Empirical Methods for the
Design of Heavily Loaded Pavement Structures
John N. MUKABI
Engineerintel - R&D, Design & Engineering, Kensetsu Kaihatsu Limited, Nairobi, KENYA
ABSTRACTThe mechanistic-empirical approach presents a new paradigm for pavement design and analyses. Nevertheless, it isimperative that the analytical functions adopted as well as the prediction equations and models employed manifest as
high a degree of precision as possible. In this Study, a unique research regime and full scale in-situ experimental testing
sections were designed and constructed with varying pavement structural configurations; and subjected to different
environmental conditions under both static and dynamic loading. The innovative methods of testing and analyticaltechniques developed and/or modified are cited. The recently developed mechanistic-empirical methods based on
results from the elaborate and comprehensive testing and research regimes designed and implemented are reported andsome case examples of their practical application for the design, construction and maintenance of heavily loaded
pavement structures are introduced.

Keywords: Pavement, mechanistic-empirical, modulus, elastic yield (limit) strain, performance prediction
1. INTRODUCTION
1.1 Overall Background
Infrastructure development is a prerequisite for theachievement of rapid socio-economic enhancement. In
particular, the development of an efficient roads network is ofparamount importance as a primary mode of transportation
for passengers, goods and services. Recent trends in East andCentral Africa depict tendencies of increased rate of rapid
development precipitated mainly by various needs due topopulation explosion. On the other hand, the lack of adequate
resources to cope with such prevailing circumstances is amajor impediment towards the achievement of any set
visionary goals.
In view of this therefore, it is increasingly necessary todevelop Value Engineering (VE) based cost-time effective
technologies that can facilitate for the comprehensive
characterization and design of pavement structures foundedon research oriented unique and innovative methods oftesting as well as sophisticated analytical techniques, versatile
enough to expound the range of extrapolation derived onlimited sets of data. Although the Mechanistic-Empirical(ME) may be a versatile application for pavement design andanalysis, the main geotechnical engineering parameters thatare employed in such a program are known to highly
influence the results of the design, analysis and performance
prediction. It is therefore imperative that the analytical
functions, equations and models adopted are of the highest
degree of accuracy and sophistication possible. The ME
pavement design approach is based on an assumption thatload-induced pavement structural responses such as stresses,
strains and deflections, can be adopted in predicting thedevelopment of pavement distress and structural deterioration,in the form of rutting and/or cracking, through the use oftransfer functions. Therefore, characterization of the linear
elastic properties is generally considered primarily important.
In this Study, however, the importance of considering theelastic properties including the elastic moduli, elastic limit(yield) strain {initial yield surface} as well as the recently
developed kinematic hardening theories [1], [2] isdemonstrated, whilst new geoscientific and geotechnicalengineering concepts related to this subject matter areproposed.
1.2 Brief Background on Some Challenges Regarding
the ME Approach
According to the Research Results Digest 307 [3], an
independent review of the MECHANISTIC-EMPIRICALPAVEMENT DESIGN GUIDE (MEPDG) prepared by theTransportation Research Board of the National Academies
within the National Cooperative Research Program of theUSA, the MEPDG uses CBR for soils and granular materials
as a basis to estimate resilient modulus (M r ). However, it
further states that research has shown that the relationshipbetween these parameters is not reliable for either material
type due to non-linear stress-strain relations and the fact that
CBR is, at best, a measure of undrained shear strength, which
does not relate closely to resilient properties at relatively low
stresses, but may relate to permanent deformation resistance.
The report further states that some of the major challenges inregard to the use of the ME design system are: a) complexities
in the calibration of the design models against actual
pavement behavior in the field due to the limitations withLong Term Pavement Performance (LTPP) database and therelated difficulties in dealing with design reliability; b)
problems encountered in developing the software; c) the
approximate nature of some of the distress prediction models;440

and, d) the need for enhanced application of models that take
into account the non-linear properties of geomaterials.
On the other hand, a review of the new MEPDG from amaterial characterization perspective by the National
Research Council (NRC) Canada [4] concludes that; although
the performance predictions of the laboratory measured
dynamic modulus are in agreement with the patterns
established in the current practice and reported in mostliterature, the following major challenges are still prevalent:a) the Asphalt Concrete (AC) dynamic modulus estimatedusing the predictive equation incorporated in the Guide
proved to be substantially different from measured values, theerror of which leads to gross underestimates of accumulated
damage resulting in undersigning the road pavement
structure; b) similarly, for unbound materials, the correlation
between physical properties (including the AASHTO
classification) and the resilient modulus, produced unreliablevalues when compared with actual measurements made in thelaboratory hence applications based on moduli estimatedusing the Guide proposed values to run the software resulted
in substantially different performance predictions comparedwith those produced when using measured modulus values.
Recently developed concepts of circumventing some of theaforementioned challenges are proposed in this paper.
1.3 Introduction of Some Advances made in this Study
Some of the most important advances made in this Study
include: a) geomathematical methods of estimating, undervarying loading and environmental conditions, modulus ofelasticity, resilient modulus, elastic limit (yield) strain, initial
yield deviator stress and strain, elastic limiting line, thelimiting elastic state, and Poisson’s ratio; all parameters ofwhich form the fundamental basis of the ME pavement
design; b) correlating NDT deflection properties to NDT and
DT elastic properties; c) developing correlations between
elastic modulus determined from mechanical tests to thatfrom geophysical tests; d) developing models forcharacterizing pavement response and performance
prediction under varying conditions; e) proposing methods ofdetermining the approximate range of dynamic and static load
effects; and, f) advancing analytical methods forcharacterizing various Geomaterials and pavement response.
2. EXPERIMENTAL TESTING AND GI REGIMESIn Ethiopia, along the heavily trafficked Addis Ababa ~Debre Markos international trunk road, twelve full-scale field
experimental testing sections were designed and constructedwith six varying pavement layers and structuralconfigurations and subjected to varying environmental
conditions over six seasonal cycles as well as different modes
and cycles of static and dynamic loading. In this case, incertain sections, mechanically stabilized Graded Crushed
Stone (GCS) and unbound lateritic gravel were adopted forthe base and subbase layers respectively [5].
On the other hand, whilst undertaking a detailed design study
for the Juba River Port access road in South Sudan, due to theheavy traffic, overloading and unconventional loading

characteristics that had been anticipated, it was imperative toinitiate a comprehensive investigation in order to realize
Value Engineering (VE) based cost-time effective design,that ensures enhanced structural performance as well [6].
The experimental testing and Geotechnical Investigation (GI)regimes adopted in this Study are reported in [7] and [8].
3. PROPOSED ME EQUATIONS FOR EVALUATION
OF EXISTING PAVEMENTS STRUCTURES
3.1 Mechanical Testing(1) Laboratory testingAt every project and in-situ experimental testing site,“undisturbed” and disturbed samples were extruded at
designated intervals and/or at points where the soil type
and/or characteristics varied; for purposes of undertaking
comprehensive laboratory testing.
Conventional and modified laboratory and field methods oftesting were adopted in this Study. The main mechanicalmethods of laboratory testing included the conventional andmodified Unconfined Compressive Strength (UCS) adopting
larger size specimens (φ ≥ 20cm), conventional Consolidated
Undrained Triaxial Compression (CUTC), Consolidated
Drained Triaxial Compression (CDTC), modified simple
shear and modified UCS dynamic loading; carried out undervarying moisture – suction and loading conditions.
In order to correct for the specimen size, some full scale plateloading tests were carried out in the field.
Basic physical and mechanical properties such as specificgravity, bulk and dry densities, Atterberg Limits, grading,moisture content and compaction were also measured andcharacterized. The data applied for modeling was determinedfrom sophisticated laboratory tests and correlated to resultsfrom field seismic and other surveys.
Details of the methods of testing are reported in [8].
The following equations, developed from Consolidation andStress Ratio (CSSR) concepts [9], were applied in correcting
for the effects of destructuration due to reconstitution.
µ eR • ηc • (q / pe ) f
 q 


=
 p 
K cfR − µ eR • Aφ • CSR R e  max
I

R

(

(

))

µ eR • η c • ( p / p e ) f
 p 


=
 p 
K cfR − µ eR • Aφ • CSR R
 e 
I

(1)

R

(

(

))

(2)

where superscripts I and R denote “intact” and“reconstituted” respectively and µR c =(q/p’) f R, η c =(q/p’) c,
and KR cf = (σ’ r /σ’ a )R ec.
(2) In-situ testingStandard Penetration Tests (SPTs) were carried out for bridgefoundation designs, while Dynamic Cone Penetration Tests(DCPTs) were undertaken for pavement structures.
The relations between the N-value and UnconfinedCompressive Strength (UCS) as well as initial stiffness
expressed as E 0 and UCS are presented in Eqs. (3) and (4)respectively.

441

qu = {[N SPT + BN ] × Aq / AN } − Bq

(3)

[

]

E0 = qu + Bq AE / Aq } − BE(4)where, N SPT is the number of blows from a Standard
Penetration Test (SPT), q u is the Unconfined Compressive
Strength (UCS), E 0 is the elastic modulus, and A N =19.6,
B N =48.4, A BC =40, B BC =98, A q =0.96, B q =2.37, A E =1024 andB E =280 are material and ground related constants.

recoverable, whereas in the latter case the response graduallytransposes from the visco-elastic towards the plastic phase.
Pd/2

hn

δ RD = [δ av. + 2V 0.5 ]× f × Lcf

t’4

δ

fδ

δ TR
δt

(5)

(6)

where, δ TR is the RRD determined at t=21°C and δ t , whichis computed from:δ t = Aδ t 3 − Bδ t 3 + C δ t − Pδ(7)where, A δ =7.2X10-5, B δ =7.6X10-3, C δ =0.27 and D δ =2.7 aredeflection-temperature related constants, while t is theaverage temperature of the asphaltic layers; t >16.5°C.
Some of the equations that were developed form thisinvestigation and applied in this Study are introduced under
section 4.2 of this paper.
3.3 Geophysical Surveys
The mathematical and regression analysis carried out withconsideration to various environmental factors yielded thefollowing equations applicable for ρ >25Ω-m.
(8)qu = Aq ln( ρ ) − Bq (MPa)

E1
E2

t3

Deflection increases

E3 with decreasing

Tt4

n
t4

δ

=

Pd

t2

where, δ av. is the average value of deflection measured undera wheel load of 5tons, V is the variance of the measurements
and f tc is the temperature correction factor expressed as:tc

Ad/2 Pd/2

t1

3.2 Deflection Testing and Measurement
The magnitude of the Representative Rebound Deflection(RRD), δ RD , for structural analysis, was computed from;
tc

Ad/2

confining stress'n σC

Fig. 1 Generalized characterization of pavement response
The definitive generalized equation delineating andsimulating the road conditions under this conceptual
framework is expressed as a function of loading conditions,pavement type (structurally), pavement layer quality,
structural thickness as well as intrinsic material properties,
inherent and induced characteristics as postulated in Eq. (10).

[v
Rc = f φ df , ∂t i , Pc , Pe , ∆te, ∆α ms

]

(10)

where, Rc represents road condition, φdf is the dynamic loadfactor, ∂ti defines the response mode factor of layer of thepavement structure, P c is the pavement configuration, P e isthe pavement layer quality, ∆t e is effective structuralv
thickness and ∆α ms=

parameter delineating moisture ~(9)suction variation.
On the other hand, the extent of deformation can be derived
by carrying out back analysis of the deformation history of an
existing pavement structure. In a generalized state, this canbe expressed as:

q u is the Unconfined Compressive Strength (UCS), E 0 is theelastic modulus, and ρ is the geo-electromagnetic resistivity.

(11)

[

]

E0 = AE ln( ρ ) − BE (MPa)

4 PROPOSED ME EQUATIONS FOR ENHANCEDPERFORMANCE PREDICTION
4.1 Generalized Pavement Response Concept
Conventionally, traffic, materials and climate are the threemain categories of input variables adopted in the MEPDG
methodology. This Study proposes a modified approach thattakes into account; traffic intensity and reciprocal
characteristics, environmental factors and the impact thereof
and unique and comprehensive geomaterial characterization.
In so doing, consideration is made of both an elastic layer
system which is insensitive to the rate of loading and thevisco-elastic layer system whereby the pavement becomessensitive to vehicle speed and frequency of loading. In theearlier case the behavior is simulated as linear-elastic andε dh = f [φ ' ,ψ ' , p'ocf , q ocf , φ 'ocf , Σf yi , δ ijo ]where, ε dh represents the parameter delineating deformation

history, φ ' is the consolidation stress ratio, ψ ' is the modifierocbetween Isotropic and Anisotropic stress paths, p 'oc
are
f ,qf
the invariant stresses under over consolidation conditions,
φ ,f is the Angle of Internal Friction within the failure zone.
4.2

Characterization of Traffic Impacted Dynamic
Loading Effects
The dynamic component of the traffic loading was simulated
as moving constant wheel loads whose transfer mechanisms
occur very slowly relative to the frequencies involved in thevertical dynamics. Computation and analysis of the stresses
transmitted is presented in [5]. In carrying out these analyses,
the mid-elements of each layer are analyzed in relation to the442applied stresses.
The dynamic traffic loading effects were analyzed on thebasis of Eqs. (12) ~ (17), which were developed in this Study.
1) Transversal propagation of stress
The modified transformed damped oscillatory equation of
motion is proposed as:[]

δ rd = co e ht sin (h 2 − ω o2 ) t + φ o
0.5

(12)

where, δ rd =rebound deflection, c o =constant representing theinitial conditions of loading, h= damping factor of thepavement structurallayer stiffness, t= response timemeasured, ω=angular frequency and 𝜙0 =constantrepresenting condition of deflection measurement.
The modified concept of energy was also applied in analyzing
the curvature of the deflection basin in relation to the elastic
moduli energy equation, which is expressed as;

(

)

[(

)

]

l w + h cos w + h t + φ 
0a
E (t ) = 0.5c02 e −2 ht 0.
5
+ f r sin 2 w02 + h 2 t + φ0

(13)
2
0

2

[(

2

2
0

)

2 0.5

]

[

]

Equations Applied for Analyzing
Impact ofEnvironmental Factors
Most tropical soils are known to be highly sensitive andsusceptible to changes in environmental factors.
Development of methods that can quantify the magnitude of
the impact of such variations on theperformance of pavement(39)
structures is therefore of great essence.
1) Effect of swelling∆ sc = ϑsc ln λsc + Β sc (%)(18)

where, ∆ sc represents swell in relation to surcharge pressure,ϑ sc = 12.9; logarithmic gradient constant for tropicalgeo-materials , λsc is the surcharge pressure in Kpa, B SC
=36.5; logarithmic intercept for most fine grained tropicalproblematic geomaterials.
2) Seasonal effects on resilient modulus (M r )The effects of seasonal changes on the resilient modulus ofsome tropical c geomaterials is presented in Eqs. (4) and (5).

ζ wMr = α gl ln M r + α gi

(19)where, ζ wMr represents the wet to dry Season resilient(41) and α= 0.0022gi

modulus (M r ) ratio, α gl

(14)

(15)

3) Moisture-suction variation
The impact of the mositure~suction variation is analysed byconsidering the correlation between(42)the change in moisture
content as a function of the plasticity ratio. Thus the resultingR

where, Δrd is the total displacement. The proposed deflective
equation of motion is then given by:resilient modulus, M r is computed from:∂ 2δ rdn∂ 2δ rdn∂ 3δ n
= Gn
+ηn
ρn
∂t 2
∂Z 2
∂Z 2 ∂t

M rR = [− Awc × ∆w c + B wc ]×(16)

where, ωη n =2G n h n , ρ= density of pavement layer, G=shear
modulus and h is damping3) Correlation between elastic modulus and deflection
Wave propagation techniques are usually used to determine
the dynamic shear and elastic moduli of in-situ ground andgeomaterials. For purposes of estimating these parametersfrom deflection testing, the following correlations weredeveloped.
𝐸 = 2[1 + 𝜐]𝑡𝑟𝛼 {𝛾𝑛 ⁄𝑔};

𝑡𝑟𝛼 𝛼𝑉𝑠2 : 𝐺 = 𝐸 ⁄2[1 + 𝜐]

= 0.54 arelogarithmic M r intercept constants for tropical geomaterials2) Shear wave propagation
Assuming the deflection at any layer n is given by,

δ rd = δ rd (Z l t ) = ∆ rd (Z )e iwt

4.3

(40)

where, f r is the force constant and l a =axle load. The energy ispostulated to decrease exponentially with the increase in time,
which is expressed in relation to the magnitude of δ rd as:

dE (t ) d 1=
l a δ rd2 + 1 f r δ 2
2
2
dt
dt

environmental factors; 𝛾𝑛 = density of layer n, 𝑔=accelerationof gravity, 𝑉𝑠2 =shear wave velocity and 𝐺=shear modulus.

(17)

where, E= Elastic (dynamic) modulus, 𝜐 =Poisson’s ratio,
𝑡𝑟𝛼 =response time between start of measurement and rebounddeflection whereby 𝛼 varies with type of pavement structure,layer
configuration,
geomaterial
properties
and

PI R
× M ri
(43)
PI
m

(20)

where, A wc = 0.092 and B wc = 1 are resilient modulus related
gradient and intercept constants, Δwc is the change inmoisture content expressed as a percentage, PI R = 16 is theReference Plasticity Index, while PI m is that determined forithe tested material and M r is the initial resilient modulus.
i

The initial resilient modulus M r can be determined from themodified quasi-empirical equation proposed as:

[ ]

3

[ ]

2

[ ]

M ri = AMr × M rAI × 10 −6 − B Mr M rAI + C Mr M rAI − DMr (21)where, A Mr =2, B Mr =0.0012, C Mr = 0.623 and D Mr = 0.775 areaverage constants determined for tropical geomaterials and

M rAI =10.3XCBR, is an equation proposed by the American

Asphalt Institute.

4434) Impact of black cotton soil ingression• Basic equations
Basic mathematical equations that can enable the quantitative
analysis of the impact of the ingression of subgrade black
cotton soil (BCS) into the upper superior subbase and basecourse layers are proposed in this Study. The ultimate CBR int.
is computed as:
32

CBR int . = ABCi ∫ − B BCi ∫ + C BCi ∫ + CBR i
BCi

BCi

BCi

(22)

where A BCi =0.057, B BCi = 2.76 and C BCi = 15.2 are BCS
inherent characteristics related constants, while ∫ BCi is the
BCS ingression content expressed in percentage terms.
• Evaluation of impact on quality of pavement layers
The deterioration of pavement structural thickness occurs
mainly as a result of cyclic action impacted by increased axle
loading, moisture ~ suction variation and intrusion ofsubgrade fines to upper layers. In order to quantify theseeffects, an Intensity factor I f , is introduced as follows:

If =

TD
xR f (1 + ∆S df x∆t s )
TEP

(23)

where, T D = design thickness, T EP = measured thickness ofthe existing pavement, R f = roughness factor expressed as

Rf = [2Ri (Rt − Ri )]0.25

given,

R i = initial roughness

value, R t = terminal roughness value, ∆S df = rate of surface
distress depreciation factor, ∆t s = time lapsed.
On the other hand, the effective structural layer thickness of
subgrade soils can be derived from the following equation.

[]t Pb = AP − B P (log CBR d ) + C P (log CBR d ) log[N / DP ] (24)2where, A P =219, B P =211, C P =58 and D P =120.

4.4
Geomaterials characterization
For the MEP design it is a prerequisite that geomaterial
properties and changes caused by loading and theenvironment be comprehensively analyzed. The basic
mechanistic properties used to evaluate the performance ofpavement geomaterials under various loading andenvironmental conditions are: 1) dynamic (elastic) modulus;2) resilient modulus and, 3) Poisson’s ratio. These parameters
are primarily adopted in the calculation of stresses, strains
and displacements within the pavement under traffic loading.
Through the application of transfer functions, the computed
results are then translated into pavement surface distresses
based on LTPP database and performance indicators
supplemented with comprehensive testing s and analyses.
The geomaterials adopted in this Study were mainly: a) black
cotton subgrade (problematic) soils; b) unbound lateritic
gravels; c) quasi-bound mechanically stabilized gravels
employing the Optimum Batching Ratio Method (OBRM) [6],[10]; d) bound graded aggregates and gravels employing the
Optimum Mechanical and Chemical Stabilization (OPMCS)

technique [6], [10]; e) geosynthetics reinforced unbound andbound materials [6]; and, f) reclaimed asphalt.
The geomaterial property input parameters that wereconsidered include; a) elastic modulus; b) shear modulus; c)resilient modulus; d) elastic limit (yield) strain {initial yieldsurface} determined within the framework of the recently
developed kinematic hardening theories [1], [2]; e) Poisson’s
ratio; f) compressive and shear strength; g) mechanicalstability due to application of OBRM;(30) h) enhanced particle
agglomeration due to effects of OPMCS; i) tensile strength; j)coefficient of thermal expansion; k) creep compliance; l)thermal conductivity; m) asphalt binder stiffness; n)aggregate properties; o) heat capacity; p)hydraulicconductivity; q) moisture ~ suction variations; r) plasticity
index; and, s) degree of layer contamination. Some of the
input parameters influenced by the variations inenvironmental conditions were computed by using Eqs. (18)
~ (24), whilst the basic mechanistic properties and other input
parameters were determined from Eqs. (25) ~ (42).
1) Moduli of elasticity and shear
Research on small strain elastic properties of mostgeomaterials within the region(31)of linear elastic andrecoverable behavior has shown that the elastic modulus isinsensitive to the rate, mode and other conditions of loading
as well as the consolidation shear stress-strain history
provided that the geomaterials are tested under post-primaryconsolidation conditions and at a constant state of stress.
However, it is the elastic yield strain limit {(ε a ) EYLS } thatvaries considerably with the changes in such conditions {alsoref. to [1], [2] and [13]}. Essentially therefore,
𝑑𝑦𝑛𝑎𝑚𝑖𝑐
Edynamic=Ecyclic=Estatic, whilst (𝜀𝑎 )𝐸𝑌𝐿𝑆> (𝜀𝑎 )𝑚𝑜𝑛𝑜𝑡𝑜𝑛𝑖𝑐.
𝐸𝑌𝐿𝑆
This fact notwithstanding, consideration was made of thevisco-elastic behavior of Hot Mix (5)Asphalt (HMA) mainly in
relation to loading and other factors such as voids contents,
effective asphalt contents, binder type and effective specificgravity of aggregates, among others.
Measurement of elastic properties within the region of verysmall strains, however, requires sophisticated equipment,
which is curtailing in terms of resources and maintenancecapability. As a consequence, Eqs. (25) and (26) whichcorrelate the initial elastic modulus, E 0 , to the maximum
deviator stress (peak strength), q max tested under CUTC
(Consolidated Undrained Triaxial Compression) conditions
using conventional apparatus, are proposed.
𝐸0 = �(𝑙𝑛7.46𝑞𝑚𝑎𝑥 + 𝐶𝜑𝐶𝑆𝑆 )⁄𝐵𝜑𝐶𝑆𝑆 � × �𝐴𝜓𝐸0 × 𝛼𝜎𝑠𝑠 � (25)𝑞𝑚𝑎𝑥 = 𝐴𝜓𝐸0 × exp[𝐵𝜑𝐶𝑆𝑆 ln{𝑚−1 × (𝜎𝑎 )𝑚𝑎𝑥 × 𝐸0 ] −
𝐶𝜑𝐶𝑆𝑆
(26)
𝐴𝜓𝐸0 = 0.134, 𝐵𝜑𝐶𝑆𝑆 = 0.418, 𝐶𝜑𝐶𝑆𝑆 0.936, 𝑎𝑛𝑑 𝑚−1 =1.32.
′
For very soft cohesive geomaterials, 𝛼𝜎𝑠𝑠 = (𝜎𝑎0)𝑚𝑎𝑥 ;
relatively stiff to hard cohesive geomaterials, 𝛼𝜎𝑠𝑠 =′{(𝜎𝑎0)𝑚𝑎𝑥 , }−1 ; for sand stone, soft rock and mudstone,′′ −1
𝛼𝜎𝑠𝑠 = {(𝜎𝑎0)𝑚𝑎𝑥 × 5 × 𝜎𝑌′ ⁄𝜎𝑎0
} ;while 𝛼𝜎𝑠𝑠 =′′−1{(𝜎𝑎0 )𝑚𝑎𝑥 × 5 × {𝜎𝑐 × 𝑂𝐶𝑅} for gravel.

444

The effects of stress states on E 0 for cohesive geomaterials
are accounted for by applying Eqs. (27) and (28) for drained
and undrained conditions respectively.
′

(27)

′

(28)

𝐸0D = 726e0.76×σa0
𝐸0U = 891e0.76×σa0

For the determination of resilient modulus, 𝑀𝑟 for unboundgeomaterials in general, Eq. (29) may be adopted with the′elastic yield limit stress, 𝜎𝐸𝑌𝐿𝑆
e determined from Eq. (30).
′2 ⁄𝑀𝑟 = 𝜎𝐸𝑌𝐿𝑆
2 𝐸0

(29)

′
= (𝜀𝑎 )𝐸𝑌𝐿𝑆 × 𝐸0
(30)
𝜎𝐸𝑌𝐿𝑆
On the other hand, E 0 can be determined from Eq. (31) for
stress-strain measured at 𝜀𝑎 =0.001% by using conventional
gap sensors, which are affordable and relatively user friendly.

𝐸𝑜 = �𝛼𝐸𝑜 (𝜀𝑎 )=0.001% + 𝛽𝐸𝑜 � × 10(𝑀𝑃𝑎)(31)where, 𝛼𝐸𝑜 and 𝛽𝐸𝑜 are determined simultaneously at
𝜀𝑎 =0.001% and 0.01%.

Prediction of shear modulus as a function of consolidationstress history and stress state can be mad from Eq. (32) below.
𝑝′𝛽[𝐺𝑜 ]𝑝′ = �𝒜𝑝𝑜′ �(𝐾𝑐𝑠 )𝛼 × � ′ � � + ℬ𝑝𝐾′𝑐𝑠 � × [𝐺𝑜 ]𝑝𝑜′𝑝𝑜𝑜2) Elastic Yield Limit Strain {(ε a ) EYLS } and Stress{ 𝝈𝑬𝒀𝑳𝑺}
The importance of determining the initial yield limit strain
and stress to a high degree of precision cannot be
overemphasized. At a designated stress state dictated by thein-situ overburden pressures, (ε a ) EYLS can be determined in %from Eq. (33), which takes into account the effects ofconsolidation and shear stress history, ℒ 𝐶𝑆𝑆𝐻 {Eq. (34)},
drainage conditions, 𝜑𝐷𝐶 {Eq. (35)}, cyclic prestraining,
𝛼𝑐𝑝 {Eq. (36)}, and loading rate, 𝜀̇𝑆𝑅 {Eq. (37)}.
𝛽𝐸𝑜 −0.1�(𝐸𝑜 )2
𝑅�𝐸𝑜 �
𝛼𝐸𝑜+ 𝑓�ℒ𝐶𝑆𝑆𝐻 , 𝜑𝐷𝐶 , 𝛼𝑐𝑝 , 𝜀̇𝑆𝑅 �

(33)

𝑡𝑝

[𝜀𝑎 ]𝜎𝑌𝑠𝑠= 𝒜ℒ𝜀 𝐶𝑆𝑆𝐻 ln(𝛿𝑆𝐶𝑇 × 𝜙𝑂𝐶𝑅 ) + �(𝜀𝑎 )𝑌𝐼 �𝑁𝐶 (34)𝐼

where 𝒜ℒ𝜀 𝐶𝑆𝑆𝐻 =1.9×10-3 is a constant, 𝛿𝑆𝐶𝑇 is thesecondaryconsolidation
factor,
𝜙𝑂𝐶𝑅
is
the
𝑡𝑝
overconsolidation factor and �(𝜀𝑎 )𝑌𝐼 �𝑁𝐶 is the initial yieldstrain determined under normally consolidated conditions at astandard time period designated after the end of primaryconsolidation.
𝐷𝐶[𝜀𝑎 ]𝜑𝑌𝐼 = �

�Δ𝜎𝑎′−2𝜈Δ𝜎𝑟′ ���Δ𝜎 ′ −Δ𝜎 ′ �𝑎𝑟𝛽�𝑑�𝑢 �𝜎′× �𝜎𝑟′ �𝑎 𝐾𝐶× �(𝜀𝑎 )𝑌𝐼 ��𝑑� � (35)𝑢where Δ𝜎𝑎′ , Δ𝜎𝑟′ are the effective axial and radial stresses
respectively determined at the threshold of �(𝜀𝑎 )𝑌𝐼 ��𝑑� � ,𝑢[𝜎𝑟′ ⁄𝜎𝑎′ ]𝐾𝐶 is the stress ratio during consolidation, 𝛽 (𝑑 )=-1,
𝛽 (𝑢)=+1 (𝑑: drained and 𝑢: undrained) and 𝜈 is the Poisson’s
ratio computed from Eq. (38).
𝛼[𝜀𝑎 ]𝑌 𝑐𝑝 = 𝒜𝜀𝛼𝑐𝑝 ln 𝛼𝑐𝑝 + �(𝜀𝑎 )𝑅𝑌𝐼 �𝐼𝜕 2 (𝜀𝑎 )𝑡𝑝

�

2
𝜕𝛼𝑐𝑝

𝜀̇

= 0 𝑎𝑛𝑑, �𝜕 3 (𝜀𝑎)𝑡𝑝
3
𝜕𝛼𝑐𝑝

𝜀̇
𝜀̇ 𝑅𝑆𝑅

�

𝛿𝑅

(𝑡𝑝)

<0

: 𝐵𝐶 =
𝜀̇[𝜀𝑎 ]𝑌𝑆𝑅= �𝒜𝜀̇ 𝑆𝑅 ln � 𝐴𝑆𝑅 � + ℬ𝜀̇ 𝑆𝑅 � × (𝜀𝑎 )𝑌𝑆𝑅
𝐼
𝐼

(36)
(37)

where, the subscripts SR denote Strain Rate, ASR: Applied
Strain Rate during testing or arbitrarily designated and RSR:
Reference Strain Rate.

(32)

where [𝐺𝑜 ]𝑝′ is the initial shear modulus at a variable stresspoint 𝑝 ′ , 𝐾𝑐𝑠 = 𝜎𝑟′ ⁄𝜎𝑎′ is the arbitrary or designatedconsolidation stress ratio traced to 𝑝 ′ , [𝐺𝑜 ]𝑝𝑜′ is the initialshear modulus determined at in-situ overburden pressure,𝒜𝑝𝑜′ = 0.95 and ℬ𝑝𝑜′ = 0.35 are geomaterial constants, thevalues of which are applicable for most natural stiff to hardclayey geomaterials, while 𝛽 =1.16 and 𝛼 =0.4 for stress statesin the 1st quadrant and 𝛼 =-1 for stress states in the 4th
quadrant accordingly.

(𝜀𝑎 )𝑌𝐼 =performed on intact specimens, field seismic surveys orotherwise computed from Eqs. (25) ~ (28).

Equation (33) can be conveniently computed from the ratio of(𝐸𝑜 )2𝑅 ⁄𝐸𝑜 , where (𝐸𝑜 )2𝑅 is measured from reconstituted
specimens, while 𝐸𝑜 is determined from CU(D)TC test3) Poisson’s ratio {𝝂𝒑𝒅 }Under partially undrained conditions 𝜈𝑝𝑑 can be determinedfrom Eq. (44) as:(𝐸 )𝜈𝑝𝑑 = �(𝐸𝑜)𝑑 (1 + 𝜈𝑢 )� − 1

(38)

𝑜 𝑢

where (𝐸𝑜 )𝑑 , (𝐸𝑜 )𝑢 are the initial drained and undrainedelastic moduli respectively.

4) Mechanical stability
Quantitative analysis of the contribution of enhanced
mechanical stability due to OBRM stabilization wasundertaken by applying Eqs. (39) and (40).
M Sf = AMS exp −1.16η

when η ≥ 0.5[

M Sf = M Sideal − AMS exp −1.16η − M Sideal

(39)

]

(40)

when 0 < η < 0.5𝑓where, 𝑀𝑆 is the Mechanical Stability (MS) factor, and forthis case, A MS is the MS constant= 178.6, η is the gradationindex=log0.01P/log(d/d max ), M S ideal=100% for non-stabilizedgeomaterial.
5) Particle agglomeration
The contribution of particle agglomeration in enhancing
strength and elastic modulus due to OPMC stabilization isquantitatively analyzed from Eqs. (41) and (42) respectively.

445

𝑖𝑞𝑚𝑎𝑥 = 𝐴𝑞 𝑙𝑛𝑎𝑓 + 𝑞𝑚𝑎𝑥

(41)

where, 𝑞𝑚𝑎𝑥 is the maximum shear strength, 𝐴𝑞 = 1.06 is amaterial constant and 𝑎𝑓 is the agglomeration factor
computed from the level of OPMC related factors.

𝐸0 = 𝐴𝐸𝑀 𝑎𝑓2 + 𝐵𝐸𝑀 𝑎𝑓 + 𝐸0𝑖(42)where, 𝐴𝐸𝑀 = 62.4 𝑎𝑛𝑑 𝐵𝐸𝑀 = 690 are geomaterial
constants. 𝐸0 is measured and/or expressed in MPa.
4.5 Modified Performance Prediction Equations
Studies on the modification of performance prediction
equations and calibration factors aimed at enhancing theirprecision, versatility, reliability and confidence levels, areon-going.
4.6 Performance Prediction Models
The performance prediction models that were adopted for thisStudy include: a) ESDAM (Environmental, Strength andDeformation Model); b) SCDRM (Structural Capacity andDeformation Resistance Model); c) CSSRM (Consolidationand Shear Stress Model); and, d) MCSTM (Modified Critical
State Theory Model).
5 ME APPROACH IN PAVEMENT STRUCTURALDESIGN, CONSTRUCTION AND MAINTENCANCE
5.1 MEP Design Approach
In this Study, the MEP (Mechanistic-Empirical Pavement)
design is incorporated into the CMD (Comprehensive
Method of Design) [11]. Conventionally, methods ofpavement design can be grouped into five categories namely:a) empirical methods; b) limiting shear failure methods; c)limiting deflection methods; d) regression methods based onpavement performance and road test; and, e)mechanistic-empirical methods [12]. In order to undertake
comprehensive design analyses for mega projects, the CMD
combines all these methods through an iterative comparative
analysis module. Basically, the MEP-CMD approach consists
1 Identification of the inputof the following steps: ○parameters through application of the equations and functions
2
introduced in this paper;○
Computation ofelastic/visco-elastic solutions under mechanical loading
3 Computation of elastic solutions under thermal
conditions; ○4 Computation of elastic solutions underloading conditions; ○5 Conversion of thephysio-chemical loading conditions; ○mechanical, thermal and physio-chemical elastic solutions tothe counterpart visco-elastic through the application of EVCP(Elastic-Visco-elastic Correspondence Principle) and the
6 Introduction of the dynamic
modules introduced herein; ○loading concept for the reciprocal loading conditions by
7
applying the BSI (Boltzman Superposition Integral); ○Analysis of integrated effects of mechanical, thermal and8 accumulation of deformations
physio-chemical loading; ○9and plastic strains; ○ Systematic introduction of otherinfluencing factors based on the SAC (Systematic Approach
10
Concept); and, ○Determination of postulated failure
mechanisms, mainly rutting, fatigue and thermal cracking
through the application of appropriate prediction models.
5.2 Construction Quality Control
The recently developed construction quality control functions
adopted for the MEP design are reported in [14].
5.3 Prediction of Maintenance Requirements
The deterioration with time of the structural capacity is
predicted by adopting Eqs. (43) ~ (45). The results for the
Juba River Port access road Case Study are plotted in Fig. 2.
−𝟏

𝒇𝒕𝒔𝒄 = 𝒇𝒊𝒔𝒄 × �𝓵𝓷𝑵𝟎.𝟔𝟐�𝒕× �𝟏 +

𝟎.𝟏𝒍𝒐𝒈𝑵𝑹𝒕
� × ℮𝟎.𝟎𝟏𝑵𝒕
𝒇𝑹
𝒔𝒄

(43)

where, 𝑓𝑠𝑐𝑅 = 𝑟𝑒𝑓𝑒𝑟𝑒𝑛𝑐𝑒 𝑠𝑡𝑟𝑢𝑐𝑡𝑢𝑟𝑎𝑙 𝑐𝑎𝑝𝑎𝑐𝑖𝑡𝑦 𝑓𝑜𝑟 𝑁𝑡 =2.2 years of post dynamic traffic loading.
Environmental factors such as moisture-suction variation due
to seasonal changes, inferior material intrusion as a result of
the combined effects of dynamic loading and water
infiltration (pumping) and land use affecting the structural
capacity and pavement structural layer thickness, arequantitatively analyzed and factored into generalized Eq. (43)by applying Eqs. (44) and (45) which incorporate the time𝑡dependant factor 𝑓𝑑𝑓
partially defining the effects of LTPP.
𝒅−𝟏

𝒇𝒕𝒅𝒇 = �℮�𝟎.𝟎𝟏𝒇𝒅𝒇𝑵𝒕� �(44)𝑑
The factor 𝑓𝑣𝑓, which defines the variation with time inenvironmental factors is computed as:

𝒇𝒅𝒗𝒇 = 𝒇𝒅𝒎𝒔𝒗 × 𝒇𝒅𝒊𝒏𝒕. × 𝒇𝒅𝒕𝒉

(45)

where, 𝒇𝒅𝒎𝒔𝒗 = moisture~suction depreciating factor, 𝒇𝒅𝒊𝒏𝒕 =BCS intrusion depreciating factor, 𝒇𝒅𝒕𝒉= pavement structurallayer thickness depreciating factor.

• Computation of Maintenance Requirement Ratio (MRR)
The computation of the MRR is made from Eq. (46) based on
the results computed and simulated by the SCDR model byadopting the T A over a loading period N t and Design Life, DL.
𝑇{𝑁𝑡=𝐷𝐿}𝑑{𝑁 =10}

𝑀𝑅𝑅 = 𝐴{𝑁𝑡=2.2} × 𝑓𝑆𝐶 𝑡
𝑇𝐴

𝑇{𝑁𝑡=0.5𝐷𝐿}

𝑁 =0.5𝐷𝐿

× 𝐴 {𝑁𝑡=2.2} × [ 𝑡 =𝐷𝐿 ] × 𝐹𝑆 (46)𝑇𝐴

𝑁𝑡

Figure 2 shows a comparison of the predicted and actual
pavement structural performance based on the results fromdeflection testing and prediction made from the SCDR modelfor three varying pavement structural configurations [6]. It
can be inferred that Type II-1 approaches the critical zone,
which is an indication of the requirement of fully-fledged
rehabilitation, after only 4.6 years; whereas Type II-3 isstructurally sound over the whole period of the Design Life.
On the other hand, based on Eq. (46) and the SCDR results,
the Maintenance Requirement Ratio (MRR), was generated
for the three varying pavement structural configurations overthe entire period of the Design Life (DL=20years). The

446

results indicated that; {Type II-1} MRR = 2.86, {Type II-2} MRR =0.88, while {Type II-3} MRR = 0.43.
1.2Structural Capacity Depreciation Factor Vs. Time
TYPE II-1 PREDICTED
Progression
TYPE II-2 PREDICTED

1.1

TYPE II-3 PREDICTEDTYPE II-2 ACTUAL
TYPE II-3 ACTUAL
TYPE II-1 ACTUALSCD Factor, fsc

1
0.9
0.8
0.7

Critical Zone

0.6
0.5
0.4

Terminal Level Line
0

2

4

6

8 10 12 14 16
Time Progression, Nt (Years)

18

20

22

Fig. 2 Progressive time dependent structural capacity
depreciation of three types of pavement structures
6 CASE EXAMPLES OF APPLICATIONThe MEP-CMD has been applied for the design of heavily
loaded pavement structures, foundations
and othergeo-structures in East and Central Africa whereby average
construction cost-time savings of approximately 40% wererealized, whilst the structural capacity and pavement
performance were appreciably enhanced [6].
7 CONCLUSIONS
The recently developed mechanistic-empirical methods
introduced in this paper form a conceptual and geoscientificbasis for geotechnical engineers in general, and pavement
designers in particular, to improve their design philosophies
geared towards VE principles, enhance reliability andprecision of prediction of overall pavement performance,
effectively simulate failure modes, quantitatively determine
impact of environmental changes, advance methods ofcharacterizing inherent as well as induced material properties,
improve construction quality control and technicallycompute maintenance requirements.
Nevertheless, the necessity of further research in this regard issignificantly apparent.
8 ACKNOWLEDGMENT

The author wishes to extend special appreciation to the Japan
International Cooperation Agency (JICA), and Kajima
Corporation, Japan, for sponsoring most of this research aswell as the Materials Testing & Research Department of the
Ministry of Roads, Kenya for their support.
Acknowledgement is also made of the relentless efforts ofmembers of the Kenya Geotechnical Society Research Team.
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[7] Mukabi J.N, “Application of Unique Methods of Testingand Analysis for the Comprehensive Design ofGeo-structuresto be published in Proc. Int. Conf. on
Geotechnical Engineering, Lahore, 2013.
[8] Mukabi JN, “Innovative laboratory and in-situ methods oftesting in geotechnical engineering, to be published in theProc. of the Int. Conf. of the Institute of Engineers Kenya ,Nairobi, 2013.
[9] Mukabi JN, “Application of Consolidation and Shear StressRatio (CSSR) concepts in foundation design and construction”
Proc. 14th ARC on SMGE Vol. II, Yaoundé, 2007,
pp.172-193.
[10] Kogi, S.K, Mukabi, J.N, Ndeda, M, Wekesa, S, “Analysis ofEnhanced Strength and Deformation Resistance of Some
Tropical Geomaterials through Application of In-situ Based
Stabilization Techniques”, Proceedings. 1st International
Conference. on Geotechnique, Environment & construction
Materials, GEOMAT, Mie, 2011.
[11] Mukabi JN, “A Comprehensive Method of Analysis for
Cost-effective Detailed Design of Pavement Structures” Proc.
14th ARC on SMGE Vol. I, Yaoundé, 2007, pp.82-112.
[12] Bhutta SA, “Mechanistic-Empirical Pavement Design
Procedures For Geosynthetically Stabilized Flexible
Pavements”, PhD Thesis in Civil Engineering, Virginia
Polytechnic Institute and State University, 1998.
[13] Mukabi J.N., “Recent Development in the Quantative
Determination of Kinematic Hardening Sub-Yield Surface
Limits”, in Procs. Int. Conf. on New Developments in Soil
Mechanics and Geotechnical Engineering, North Cyprus,
2012.
[14] Mukabi JN, Kimura Y, Murunga PA, Njoroge BN,
Wambugu J, Sidai V, Onacha K., Kotheki S, Ngigi A, “Caseexample of design and construction within problematicsoils”, in Procs. Int. Geotechnical Conference on
Geotechnical Challenges in Megacities, Geomos, Moscow,
2010, vol. 2, pp 1172-1179.

[1] Mukabi JN, “Characterization and modeling of various
aspects of pre-failure deformation of clayey Geomaterials –447

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Design of Support System for Excavation in Black Cotton Soils in Guntur,
India
D. Neelima Satyam1, Akhila Manne2

Earthquake Engineering Research Centre,
International Institute of Information Technology Hyderabad,
Gachibowli, Hyderabad-32, India.
ABSTRACT
Deep excavations are one of the venturesome tasks taken up by geotechnical engineers. Adept and secure design ofdeep excavations in especially clayey strata is a defying task per se it involves assumptions. Controlling the mobility ofclayey soil is intricate. Deformations are consequential to stress release and redistribution in deep excavations andstability controls the design of the support systems. Usually, it is the supervision of deep excavations (Burland andHancock 1977, O'Rourke 1981; Finno et al. 1989, Hansmire et al. 1989, Ulrich 1989; Whitman et al., 1991; Ikuta et al.,1994) at field that succor to attain economy and safety. It also aids in prediction of the ground settlement on adjacentstructures. The objective of this paper is to discuss the design of sheet pile, supporting structures and challenges faced in
the construction of a sump in clayey soil site. The case of study is located at Guntur district in the state of Andhra
Pradesh, India. Standing on a bed of black cotton and plastic clays with water table as low as 0.5m, the site renders ageotechnical challenge.
Keywords: Deep excavation, Sheet pile, Finite Element Analysis, PLAXIS[1]INTRODUCTION

Construction of deep basements, subways and service tunnels
require deep excavations eventually making them contingent
in the construction activities. Excavation is one of the mosthazardous construction operations as the behavior ismultifaceted and their failures are rapid. The effects arecaused due to the decrease in vertical stress and loss in lateralsupport and hence requires the examination of field
performance and monitoring [1]-[3]. Stability anddeformation are the weightage factors in the performance of adeep foundation. If the factor of safety is small, strains are
small and inturn the ground movements are small. The
deformations damage the adjacent structures like buildings
and utilities. Severity of this damage depends on the pattern
and movements around the excavation.
Soil type dictates the design and performance of suchexcavations [4]-[6]. Including soil properties, depth of
excavation and workmanship are intrinsic in controlling theearth movements [4].To verify the performance of a deep
excavation, their analysis is required. Though stability is easy
to analyse using equilibrium calculations, deformations are
difficult to predict and require finite element analysis.
References [7]-[11] used 3D finite element analysis to study
deep excavations. Study and analysis done by [10] and [11]have shown that a significant reduction in deformation in thecorners was observed when corner restraints were used whencompared with that for a plane strain condition. Fig 1

illustrates the effect of support system stiffness on the wall
displacements.
In this paper the design of sheet piles as temporary support
system for excavation in deep clay is dealt .The excavation
has a 17m x 13.26 m plan area for construction of anunderground sump with 6 m excavation depth at the Reliancecash and carry construction site in Guntur located in the stateof Andhra Pradesh, India.
[2]

SITE DETAILSThe site is blanketed with plastic silty clays up to 7.5m below
the ground surface, underlain with about 5m of sandy clays(Fig). In order to excavate to the required depth of about 6 m
total is a major challenge of controlling the soil collapse as
well as the lateral support of the excavation.
In detail, the ground consists of 3 layers (Fig. 1). The upper
most layer is 5.0 to 7.5 m thick silty clay (layer I). Beneath thefill is sandy clay (3.5 to 4.5 m) thick layer (layer II) and it isunderlain by greenish white rock fragments / reddish clayey
sand gravel (11m to 13.5m, layer III) .The recorded SPT ‘ N’
value ranges from 5-36 in the topmost layer ,19-61 for the
sandy clay layer and 41-99 for the rock fragments in the thirdlayer. The groundwater level was located at about 0.5m below
the ground surface during the exploration in 2011(May) but issuspected to be seepage. The results of a number of field andlaboratory soil tests are summarized in Fig. 2 and Fig. 3. The
natural water content varies between 37 and 7% (including
the pebbles) and shows a general decreasing trend with depth.

448

[3]

MODELING AND ANALYSIS

With the increase in the complexity of a geotechnical problem
it is preferable to use a numerical model. Partial differential
equations are employed with a combination of different
variables and the calculations are used to mould the model.
The variables maybe material parameters, stresses, andstrains etc, and have a complex correlation. According to[12] the maximum horizontal wall displacement (δhm) lies
between 0.5% H and 1.0 % H, on average at 0.87% H (Fig.3).

Figure 1 Chart for estimating maximum lateral wall movements andground surface settlements for support system in clays [6].

The use of the sheet piles as temporary support system for thecurrent problem of study satisfies the limits imposed by thedesign specifications and also addresses the constructabilityof the underground water tank. Finite element analysis isused, which is vital not only in the evaluation of the behavior
and design of the support system for underground sump, butalso in the evaluation of its impact on adjacent structures. The
analysis is primarily based on the theory of earth pressure
developed by Rankine, 1857. When designing sheet pilewalls, the partial coefficient method is used, both inserviceability limit state and ultimate limit state.
The finite element software PLAXIS [13] has been used tomodel the staged excavation of the underground sump. Fig
4,5 show the excavation pit section details and plan withstructural support system.

Figure 2 Predominant soil profile with SPT ‘N’ value

Table 1 Soil Properties
Depth
(m)

Soil Type and IS Classification

0-2.5

Plastic silty clay [MH-CH]

2.5-4.5
4.5-7.5
7.5-9.
9-12
12-15
15-20

Silty clay [MH-CH]
Plastic silty clay [MH-CH]
Sandy clay +lime pebbles
Sandy clay +lime pebbles
Clayey sandy +pebbles [MI-CI]
Rock fragments [SC]SPT ‘N’
8
11
15-19
30
32-38
46-47
54-66

The liquid limit in the upper part is around 40 % which can beclassified as plastic silty .Unconfined compressive strengthand direct shear tests were conducted during the exploration
phase.

449

Figure 3 Variation of maximum horizontal displacements withexcavation depth [12]

properties from the geotechnical investigation reports wereused. Struts or braces are modeled as bar elements whichhave only axial forces and no resistance to bending. The
connection between the wall and the strut is consequently
analogous to a pin connection. Pre-loads are specified as partof the strut definition. ISMB 250 is used at two preferable
levels at 1.0 and 4.0m below the EGL.
[4] RESULTS AND CONCLUSIONS
Understanding of the load transfer mechanism around theconstruction site, superior workmanship, soil properties
information is required for deep excavations. Prediction of
accurate design before the construction is necessary,
especially when clay soils are being dealt. Catastrophic
failures of the excavation and structural failure can be
avoided by detailed deformation analysis and site inspection
during construction process.

Figure 4 Excavation section with support details

In the current case, a temporary support system using sheet
piles TSP III with 400 x 125 x 13mm has been adopted foradequacy of the lateral loads. Fig.7 details the deformed meshof the sump after the analysis. The total displacements andvertical displacements at 14.6mm and horizontal
displacements at 10.58mm are shown in Figs 8-10.

Figure 5 Excavation Plan details with support system

Figure 7 Deformed mesh of sump

Figure 6 Geometry of the model with TSP III sheet pile for 12m

The geometry of the sheet pile used is TSP III with 400 x 125
x 13mm section piles (Fig. 6) with unit weight of 60kg/m
from Hanwa Singapore is provided. In addition, soil
450

Figure 8 Total displacement (14.66 mm)

Finally, sheet piling has been designed to facilitate the
excavation in the site blanketed with clay as sheet piling also
does not require much expertise and time. However, skill isnecessary to safely install walers and toms which support thepiling.
[5] REFERENCES

Figure 9 Vertical displacements (14.66 mm)

Figure 10 Horizontal displacements (10.58 mm)

Figure 11 Effective stress (145.50 kN/m2)

Figure 12 Total stress (208.66 kN/m2)

The values of total and effective stress are derived to be
145.5kN/m2 and 208.6 kN/m2 (Fig.11, 12)451[1] Burland, J.B., and Hancock,
R.J.R. (1977).
Undergroundcarparkat the Houseof
Commons, London: Geotechnical aspects. The
Structural Engineer, No. 2, Volume 55, 87-100.
[2] O'Rourke, T.D. (1981). Ground movements causedby braced excavations. J. Geotech. Enging Div. Am.
Soc. Civ. Engrs 107, GT9, 1159-1178.
[3] Finno, R.J., Atmatzidis, D.K. and Perkins, S.B.
(1989). Observed performance of a deep excavationin clay. J. Geotech. Engrg., ASCE, Vol.115, No.8,
pp 1045-1064[4] Peck, R.B. [1969]. “Deep excavations and tunnelingin soft ground. State of the Art Report” Proc. Of the
VII ICSMFE, Mexico, pp. 225-290.
[5] Goldberg, D.T., Jaworski, W.E., and Gordon, M.D.
(1976). “Lateral support systems and underpinning”.
Rep. FHWA-RD-75-128, Federal Highway
Administration, Washington D.C.
[6] Clough, G.W. and O'Rourke, T. D., (1990).
“Construction induced movements of in situ
Walls”. In Proc. Design and performance of earth
retaining structure, ASCE Special conference,
Ithaca, New York, pp 439-470.
[7] Borja,R.I.
(1990).”Analysis of incremental
excavation based on critical state theory”, Journal of
Geotechnical Engineering, ASCE, 116(6), 964-985[8] Schweiger, H.F. and Freiseder, M (1994) “Three
dimensional finite element analysis of diaphragmwall construction” Proc. Of the Eighth Int.Conf.on
Computer Methods and Advances in Geomechanics,
8(3),2493-2498.
[9] Tabrizi,K.S.,Gucunski,N.,and Maher, M.H.(1995).
“3-D analysis of excavation of a soil-nailwall,”,Second congress on computing in Civil
Engineering, ASCE, 1,812-819.
[10] Ou, C.Y. and Chiou, D.C. and Wu, T.S. (1996).
“Three dimensional finite element analysis of deep
excavations”. Journal of geotechnical engineering,
Vol 122, No.5, May,1996, pp337-345.
[11] Lee,F.F.,
Yong,K.W.,Quan,K.C.N.,and
CheeK.T.(1998). “Effect of corners in strutted
excavations:field monitoring and case histories”,
Journal of Geotechnical and Geoenvironmental
Engineering,ASCE,124(4);339-349[12] Moormann, C. (2004). “Analysis of wall and groundmovement due to deep excavation in soft soil basedon a new worldwide database”, Soils andFoundations, Vol. 44, No. 1, 87-98.
[13] Brinkgreve R.B.J. 2002. PLAXIS: Finite element
code for soil and rock analyses, Balkema

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Numerical Modeling Of Pullout Capacity Of A Suction Pile In Sand Under
Oblique Load
Md. Iftekharuzzaman1 and Dr. Bipul Hawlader 2
Memorial University, St. John's, Newfoundland, Canada

ABSTRACT
Suction piles are widely used in mooring system for deep water oil and gas development projects. In this study
three-dimensional finite element analyses are performed to estimate the pull-out capacity of a suction pile subjected to
oblique loading. The numerical modeling is performed using ABAQUS finite element software. The effects of two key
variables, loading angle and mooring line position, are investigated. The finite element results are compared withcentrifuge test results available in the literature. The maximum pull-out capacity is obtained when the mooring line is
attached at approximately 75% depth of the pile for the cases analyzed in this study.
Keywords: Suction pile, oblique load, finite element analysis, pull-out capacity.

1. INTRODUCTION
A suction pile (also known as suction caisson) is a large
cylinder, usually made of steel, with an open bottom and aclosed top that is installed in the ground mainly by suction
applied by pumping water out of the caisson interior. Suction
piles have been widely used in offshore industries ranging
from anchor for floating facilities to offshore foundations.
Geometrically the suction piles are larger in diameter thantypical piles used for foundation [1]. Fig. 1 shows some use ofsuction piles for various projects in the world. Suction piles
could be installed both in clay and sand sea beds, although themechanism during installation is different. Houlsby andByrne (2005a, b) [2], [3] present the design procedure for
installation of suction piles in sand, clay and othergeomaterials.
Suction piles are widely used in mooring system for deep
water oil and gas development projects, where the pull-out
capacity is one of main requirements. The piles are normallypulled by a chain connected to the pad eye on the side of thepile. The inclined pull-out capacity of suction pile depends on
both horizontal and vertical load capacity.
Yoshimi (1964) studied the behavior of rigid vertical andbatter piles in a cohesionless soil subjected to inclined loading[4]. Broms (1965) also analyzed the Yoshimi (1964)
experiments and proposed an equation for the pull out
resistance [5]. Poulos and Davis (1980), based on theexperiments of Yoshimi (1964) and the analysis of Broms(1965), proposed a simplified theoretical method to predictthe ultimate resistance of vertical pile under oblique loading[6].
Finite element analysis of a single pile under lateral andoblique pulling has also been conducted by some researchers.

Erbrich (1994) conducted a series of finite element analysis toestimate the capacity of suction caissons used as foundations
for fixed offshore steel platforms [7].

Fig. 1. Suction piles used in various projects (Byrne 2005b)
Bye et al. (1995) presented the design analyses of theEurope 16/11E and sleipner T foundation in dense sand [8].
Sukumarn et al. (1999) and Sukumaran and McCarron (1999)
showed the application of the finite element method toestimate the capacity of suction pile foundations installed insoft clays and subjected to axial and lateral loads under
undrained conditions [9], [10]. Handayanu et al. (1999 and2000) used a quasi-three-dimensional finite element model tostudy the response of suction caissons subjected to vertical452

uplift and inclined loads [11]. Deng and Carter (2000)
presented finite element analyses using axisymmetricelements and proposed a simplified relationship for
estimating inclined pull-out capacity under drained condition[12]. Zdravkovic et al. (2001) conducted finite element
analyses to study effects of load inclination, caisson aspectratio, soil adhesion, and soil anisotropy on behavior of suction
piles [13]. Cho and Bang (2002) examined the application of
the failure envelop developed by Bransby and Randolph(1999) from the observation in clay for estimating inclined
load capacity of a suction pile in sand [14]. Bang et al. (2011)
conducted a series of centrifuge tests to estimate the pullout
capacity of a suction pile installed in sand [15].
This paper presents three-dimensional finite element
analysis of a steel suction pile embedded in sand subjected to
oblique loading at different load inclination and mooring
positions. A total of 25 cases are analyzed to evaluate thepull-out capacity of a suction pile. The finite element results
are compared with centrifuge test results. The effects ofloading angle and mooring positions on ultimate pull-out
capacity, lateral displacement and soil reactions on suction
piles are discussed.

perpendicular to it, while the other three vertical sides of thesoil domain are restrained against lateral displacement usingroller supports at the nodes. The bottom boundary is retrained
from any vertical displacement, and the top boundary is free
to displace.
The finite element mesh used in this study is shown in Fig.
3. The elements used are the solid homogeneous C3D8R
element, which is an 8-noded linear brick, multi-material andreduced integration with hourglass control.

6m

2. PROBLEM DEFINITION
Finite element (FE) analyses are performed to calculate thepullout capacity of suction piles. A suction pile of diameter D
and length L installed in sand is loaded for different mooring
positions and loading angles. The notations used in this study
are shown in Fig. 2. The load is applied at five pad-eye
locations as shown by solid circles on the left.

θD

Fig. 3. Finite element model with medium dense mesh
The numerical analysis consists of two major steps:
gravity and loading step. In gravity step the soil domain isloaded up to in-situ stress condition, and in the loading steppreselected lateral and oblique displacements are applied on
the nodes at desired depth (pad-eye location) on the left sideof the outer surface of the pile as shown in Fig. 2 and Fig. 3.
3.1 Pile Modelling
A steel pipe pile of 6 m length and 3 m diameter with 100 mm
wall thickness is modeled in this study. The pile is modeled as
an elastic material with modulus of elasticity (E p ) of 208×106
kN/m2 and Poisson’s ratio (ν p ) of 0.3.

L

Pad-eyeFig. 2. Problem definition
3. NUMERICAL MODELLING
In this study, numerical analyses are carried out using finite
element software ABAQUS/standard 6.10-EF-1. A cuboid
soil domain of 40 m length, 20 m width and 20 m height asshown in Fig. 3 is modeled. The size of the soil domain issufficiently large compared to the size of the pile andtherefore boundary effects are not expected on calculatedload, displacement and deformation mechanism. The verticalplane of symmetry is restrained from any displacement

3.2 Soil Modelling
The sand is modeled by the Mohr-Coulomb constitutive
model available in ABAQUS FE software using thefollowing soil parameters: angle of internal friction, φ′=39°;
dilation angle, ψ = 9°; modulus of elasticity, E s = 60,000 kPa;
and Poisson’s ratio, ν p =0.3.
The location of the groundwater table is at the ground surface.
Submerged unit weight of 8.2 kN/m3 is used for soil layers.
Geometry and mechanical properties used in the analysis areshown in Table 1. Note that the geometry and soil parameters
mentioned above and in Table 1 are very similar to Bang et al.
(2011) as the numerical results presented in this study areverified using their test results [15].
The soil/pile interaction has been modeled using Coulomb
friction model, which defines the friction coefficient (µ) asµ=tan(φ µ ), where φ µ is the pile/soil interface friction angle.

453

The value of φ µ is assumed to be equal to 0.7φ′ in thisanalysis.
Table 1. Geometry and mechanical properties used in theanalysis
Pile:
Length of the pile (L)6m
Diameter of the pile (D)3mWall thickness of the pile (t)100 mmModulus of elasticity of pile (E p ) 208x106 kN/m2
Poisson’s ratio (ν p )Soil (sand)
Modulus of elasticity, E s
60,000 kN/m2
0.3
Poisson’s ratio, ν s
3Submerged unit weight of soil, γ′ 8.2 kN/m
39°
Angle of internal friction, φ′ p9°Dilation angle, ψ
3.3 Mesh Sensitivity Analysis
The size of the mesh has a significant effect on finite element
modeling. Often finer mesh yields more accurate results but
computational time is higher. For successful modeling ofload-displacement behaviour of piles under oblique loaddenser mesh should be used near the pile. As shown in Fig. 4that smaller soil elements are used near the pile and the size ofthe elements are increased with radial distance from thecentre of the pile. Also denser mesh is used in the top 6 m ofsoil where the pile is located. Below 6 m depth coarser mesh
is used which does not have significant effect on calculation.

Fig. 4. Mesh sensitivity analysis
After several trial analyses with different mesh size, theoptimum mesh is selected. Fig. 4 shows the lateral load
versus lateral displacement of the pile for three different typesof mesh. In the coarse mesh, a total of 5,380 elements, in themedium dense mesh 7,300 elements and in the fine mesh
11,140 elements are used. The distribution of mesh size isshown in Fig. 3. As shown in Fig. 4 that the number of
elements has considerable effect on force-displacement
behavior. As for very fine mesh the computational time ishigh, the analyses presented in the following sections are
conducted using medium mesh with 7,300 elements.
3.4 Centrifuge Modeling
Bang et al. (2011) conducted a series of centrifuge model testof a suction pile embedded in sand to evaluate its inclined
pull-out capacity [15]. The effects of load inclination angle
and the point of mooring line attachment are studied. The
tests were conducted at 100g using geotechnical centrifuge.
The test condition and geotechnical properties of sand aresimilar to those described in Section 3.1 and 3.2.
4 NUMERICAL RESULTS
Numerical modeling is carried out for a single pile applying
the load at five mooring positions: 5%, 25%, 50%, 75% and
95% distance from top of the pile. At each mooring position
the load is also applied at five different angles of inclination
with the horizontal axes: 0°, 22.5°, 45°, 67.5° and 90°. Thatmeans, a total of 25 (5×5) numerical modeling are conducted
to show the effects of inclination angle and mooring position
on pull-out capacity.
4.1 Load-Displacement curves
The variation of total load with total displacement of thesuction pile for loading at 5% mooring position is shown in
Fig. 5. The total load is calculated from the vertical andhorizontal nodal force components at the point of loading.
Bang et al. (2011) reported the pull-out capacity for similar
condition from of number of centrifuge tests. The range of
their pull-out capacity is shown by the arrows on the rightvertical axis. As shown the numerical prediction in thepresent analyses reasonably match with the centrifuge test
results [15].
Various methods are available in the literature to estimate
the pull-out capacity of a pile. In this study, the pull-out
capacity is determined by the first peak load inload-displacement curve. In the absence of a clear peak, theload corresponding to the displacement of 10% pile diameter(i.e. 0.1× 3 =0.3 m) is considered as pull-out capacity of thepile [16]. The pull-out capacity is shown by the verticalarrows in Fig.5.
Fig. 6 shows the variation of the lateral load with lateraldisplacement for five different mooring positions. Again, thepull-out capacity obtained in centrifuge tests [15] is shown on
the right vertical axis using horizontal arrows. The pull-out
capacity (vertical arrows) in the present finite element
analysis compares well with centrifuge test results. The
pull-out capacity increases as the depth of mooring position
increases. The maximum pull-out capacity is obtained for
75% mooring position. After that, for example at 95%
mooring position, the pull-out capacity decreases. This isbecause of the displacement and rotation of the pile undermooring force, which will be further discussed in thefollowing sections.

454

Fig. 5. Total load vs. total displacement for 5% mooring
positionFig. 7. Pull-out capacity for different loading angle andmooring position
4.3 Plastic Strain and Displacement Vector Diagram
Fig. 8 and Fig. 9 show the plastic strain and displacement
vectors (total) for two cases. For 5% mooring position theplastic strain mainly developed on the left side of the pile.
The bottom of the failed soil wedge on the left is almost
linearly extent to the bottom of the pile. On the other hand theshape of the plastic zone on the left side of the pile for 75%
mooring position (Fig. 9) is different from the shape shown in
Fig. 8. This is again because of rotation of the pile. As thedeformation pattern is different, it should be considered inestimation of pull-out capacity.

Fig. 6. Lateral load vs. lateral displacement for different
mooring positions.
4.2 Pull-out Capacity
As mentioned before that a total of 25 finite element analyses
are presented in this paper. The load-displacement curves
obtained are very similar to curves presented in Figs. 5 and 6.
The pull-out capacity obtained for various conditions isplotted against load inclination angle for different mooring
positions in Fig. 7. As shown, at 75% mooring position the
maximum pull-out capacity is obtained. It is to be noted herethat the pull-out capacity at zero inclination angle (horizontal
loading) is highly dependent on mooring position. However,for inclination angle of 90° (vertical loading) the pull-out
capacity is almost independent of mooring position.

Fig. 8. Maximum principle plastic strain and displacement
vector diagram for 5% mooring position and 0.5m
displacement at 22.5 degree angle

Fig. 9. Maximum principle plastic strain and displacement
vector diagram for 75% mooring position and 0.5m
displacement at 15 degree angle.
455

rotated toward the right when it is at 95%. The minimum
rotation is occurred for 75% mooring position.

4.4 Lateral Displacement vs. Depth
Fig. 10 shows the variation of lateral displacement along thecenterline of the caisson against depth for different loading
angle at 5% mooring position at 0.3 m displacement. The
lateral displacement is obtained from the displacement of thecenter nodes of the pile. As shown in this figure that thelateral displacement is almost linear. That means this shortpile is rotated almost as a rigid body. The degree of rotation
and also the center of rotation is dependent upon the angle ofloading. The rotation has a significant effect on pull-out
capacity.

4.5 Mobilized Soil Reaction
Fig. 12 shows the variation of mobilized soil reaction (loadper unit length of the pile) with depth at 0.3 m displacementfor different loading angle at 5% mooring position. In finite
element analysis, the soil reaction is obtained from the sum of
the lateral component of nodal force at a particular depth
dividing by the vertical distance between two node sets at the
point of interest. As noted, with increase in loading angle thesoil reaction decreases because of less lateral deformation andinteraction between horizontal and vertical movement.

Fig. 10. Lateral displacement for different loading angle at
5% mooring position

Fig. 12. Soil reaction for different loading angle at 5%
mooring position

Fig. 11. Lateral displacement for different mooring positions

Fig. 13. Soil reaction for different mooring positions

Fig. 11 shows the variation of lateral displacement withdepth for different mooring position under lateral loading for
0.3 m lateral displacement. The suction pile is rotated toward
the left when the mooring position is less than 50% while it

The effect of mooring position on mobilized soil reaction
is shown in Fig. 13. As shown the mobilized soil reaction isdependent on mooring position. For mooring position lessthan 50% the pattern of mobilized soil reaction curve is456

similar while after 50% mooring position it is different, which
is because of the rotation of pile.
5 CONCLUSION
A total of 25 finite element analyses are conducted to evaluate
the pull-out capacity of a suction pile. The effects of two key
variables examined in this study are: loading angle andmooring position. Finite element results have been comparedwith centrifuge test results. It is shown that the pull-out
capacity of suction piles increases as the mooring position
moves towards the pile tip and the 75% mooring line
attachment gives the maximum pull-out capacity for the cases
presented in this paper. Pull-out capacity also decreases withincrease in loading angle. The shape of the soil failure wedge
is dependent on mooring position and loading angle which
has a significant effect on pull-out capacity.
6 ACKNOWLEDGMENT
The authors are grateful to the technical and financial
supports provided by MITACS, PRAC and NSERC
Discovery grant.
7

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Oxford.
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158, No 2, pp. 75-82.
[3] Houlsby, G.T. and Byrne, B.W., “Calculation proceduresfor installation of suction caissons in sand,” Proc ICE Geotechnical Engineering, 2005a, Vol. 158, No 3, pp.
135-144.
[4] Yoshimi, Y., “Piles in cohesionless soil subject tooblique pull,” Journal of the Soil Mechanics andFoundation Division, ASCE, 1964, Vol. 90, pp. 11-24.
[5] Broms, B. B., “Piles in cohesionless soil subject tooblique pull,” Journal of the Soil Mechanics andFoundation Division, ASCE, 1965, Vol. 91, pp. 199-205.
[6] Poulos and Davis, “Pile foundation analysis and design,”
1980, New York: John Wiley & Sons.
[7] Erbrich, C.T., “Modeling of a novel foundation foroffshore structures,” Proceedings of the 9th UK
ABAQUS User's Conference, 1994, Oxford, England,
pp. 235-251.
[8] Bye, A., Erbrich, C., Rognlien, B., and Tjelta, T.I.,“Geotechnical design of bucket foundations,” Proc. of
Offshore Technology Conference, 1995, OTC 7793.
[9] Sukumaran, B., McCarron, W.O., Jeanjean, P., andAbouseeda, H., “Efficient finite element techniques forlimit analysis of suction caissons under lateral loads,”
Computers and Geotechnics, 1999a, Vol. 24, pp. 89-107.
[10] Sukumaran, B., and McCarron, W.O., “Total andeffective stress analysis of suction caissons for Gulf of
Mexico conditions,” In Analysis, Design, Construction

and Testing of Deep Foundations, Edited by J.M.
Roesset, Proceedings of the OTRC'99 Conference,
1999b, Geotechnical special publication No. 88, pp.
247-260.
[11] Handayanu, Swamidas, A.S.J., and Booton, M.,“Behavior of tension foundation for offshore structures
under extreme pull-out loads," Proceedings of 18th
International Conference on Offshore Mechanics andArctic Engineering, St. Johns, Newfoundland, Canada,
1999, OMAE99/OFT-4204, Pp. 635-641.
[12] Deng, W., and Carter, J.P., “Inclined uplift capacity ofsuction caissons in sand,” Offshore Technology
Conference, 2000a, Houston, Texas, OTC 12196, pp.
809-820.
[13] Zdravkovic, L., Potts, D.M., and Jardine, R.J., “Aparametric study of the pull-out capacity of bucket
foundations in soft clay," Geotechnique, 2001, Vol. 51,
No. 1, pp. 55-67.
[14] Cho, Y. and Bang, S., “ Inclined Loading Capacity ofSuction Piles,” Proceedings of The Twelfth International
Offshore and Polar Engineering Conference,
Kitakyushu, Japan, 2002, May 26–31, pp. 827-832.
[15] Bang, S., Jones, K. D., Kim, K. O., Kim, Y.S. and Cho,
Y., “Inclined loading capacity of suction piles in sand,”
Journal of Ocean Engineering, 2011, Vol. 38, pp.
951-924.
[16] Hesar, M., “Behaviour of pile-anchors subjected tomonotonic and cyclic loading,” 1989, PhD Thesis,
University of Sunderland, UK.

457

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051“Slope Failure in Residual Soils of Peru”
A.Carrillo-Gil , P.E., MSc., DrEng., F.ASCE
Professor Ricardo Palma University, Lima, Peru.
A. Carrillo-Acevedo P.E., MSc., M. ASCE
Member of the Board AC Consulting Engineers, Lima, Peru.
ABSTRACTThe main objective of this paper is to present a real case occurred in residual soils from the Peruvian Amazon plane, inorder to show the positive effects of the stabilization with drainage and the pore pressures dissipation that previously
had originated large landslides in the season of the annual water level decrease of the Amazon River. This happens in avery short time and decreases around 12 meters in a fast way. This effect decreases the shearing strength of thesaprolitic soil underlying, producing instability in its banks and important damages in the works of civil engineeringover the surface. The results of the practiced instrumentation allowed a better planing and distribution of the drains in
the affected area as well as an interpretation of the registered movement with biaxial inclinometers and the waterpressures with pneumatic piezometers. All of them were associated with the extensive rains of the area, the movementof the riverbed and the rapid drawdown of the water, minimizing the risk and creating better possibilities for future
investments.
Keywords: residual soils, slide stabilization
1. INTRODUCTION
The stability of the riverbanks in the Peruvian Amazon jungle
present a great number of technical problems not existing inother places, since in very few regions of the world arepresent the atmospherical, environmental or hydrologicalconditions that prevail in this region, adding to these factors
the lack of conventional construction materials. The erosion
and sedimentation phenomena that alternatively occur in bothmargins of the Amazon river, and the continuous coursechanges between the subsequent years, present additional
problems and large challenges to the application of theknowledge of the geotechnical engineering.
To offer some explanation to the movement of the meandersof the Amazon river, the reasons are: Soil with very low
gradient and smootly sloped toward to the East, in the order of
1: 20,000, that offers greater or smaller resistance to the water
flow. The changes of water level between flood and ebb times,that reach fluctuations from 10 to 12 meters.
The tectonic movements in the Amazon zone are small,
however the surface of the land bark suffers level changes,
originating possible displacement in the bed of the rivers.
According to what is shown previously, the Amazon river has
impacted strongly on the riverbank causing considerable
landslides (Fig. 1), being produced phenomena of instability.
Phenomena go advancing downstream initially as erosion to
end afterwards as sedimentation and therefore stabilization of
the slide critical area [1].
During more than 40 years they have been producing
landslides that have considerably damaged different types of458

engineering works placed in the banks of the Amazon river inthe region of Peru, when the river impacts directly on thecritical border, and increasing gradually according to the riveris going far.

Fig, 1. Landslide at the Amazon Riverbanks.
2. GEOLOGYCAL AND GEOTECHNICAL SETTING
The general geology considers that a large part of the Amazon
region has stayed covered during the interglacial periods ofthe quaternary by an interior sea of shallow water when thelevel of the oceans had 100 meters above of the existing now(330,000 years ago) it also to fluctuate during several glacial
and interglacial periods forming terraces throughout the water
courses, dropping to 100 meters below of the original level
during the last Glacial Era (17,000 years ago) and remaining
in these deep channels the large rivers, between them theAmazon river, raising afterwards to the current level (6,000
years ago).
The accomplished studies establish that in the high jungle andin the limits of the low jungle are found so much igneousrocks as sedimentary, while in the low jungle prevail
saprolitic soils originated by the sedimentary rocks of the
terciary and quaternary and they are formed mainly bysandstones, shales and clays. The general description of thegeomorphology of the Amazon region indicates that the low
jungle is substantially flat and as said remain, its height varies
between 80 to 400 meters above mean sea level. Due to thissmall difference of elevation the rivers flow slowly, getting in
the dry station the appearance of lakes. This region of the
Amazon plain, can be indicated as advanced erosion type (Fig,2). The Amazon plain is characterize by its great humidity
and soil covered by a dense tropical vegetation.

Fig. 2. Erosion Consequences at the Amazon River.

3. AMAZON RIVER MEANDERS
Thre Amazon River is formed from the confluences of the
Marañón y Ucayali Rivers nearby city of Nauta. This islocated 74 miles from the city of Iquitos. It has a length of
3,762 kilometers (2,031.3 nautic miles). 570 Km (307.7
miles) correspond to Peruvian territory. Its course ispredominantly to the east to the Atlantic Ocean with wide
meanders and numerous islands; its stream bed varies from
1,000 and 4,000 meters in the Peruvian territory with low
river banks that have floodings in high water level seasons.
Human settlements are located usually in high river banks
areas. River bed is formed with muddy material and finesands. Its main affluents are the Napo River, Putumayo Riverby the left side, Yavari by the right side to 41, 487, and 270
miles respectively to the city of Iquitos.
The river rise happens from November to May. Its peakoccurs in April and May. Drawdown happens in the second
half of the month of May until September. Maximum
drawdown occurs in August and September. Drawdown
happens faster than the rise.
The Amazon River does not present a define bed. On contrary,
it flows creating meanders over a tropical soft soil potentiallyvariable. The meanders move downstream, originating
erosion and sedimentation. Comparing information from1948 to 2009, it is observed great erosion that allowed themeanders advance upstream but no at the velocity that theriver normally used to, causing its deformation upstream anddownstream and creating landslides ant the river banks slopes[2]-[4].
According to this, the Amazon River has strongly impacted
its stream bed, causing deformation of its meanders at the
upstream area as well, modifying its stream and creating
diverse problems related to slope stability. The first registeredproblem as a landslide was observed in June 1992.
Afterwards, large landslides happened at the critical seasonsof discharge in 1993, to collapse in its entirety in 1994.

MAY 2009

LEVEL OF THE RIVER (m.a.s.l.)

CURRENT RISELEVEL OF THE RIVERHIGHER LEVELAVERAGE ANNUAL
MOVEMENT OF THE RIVERJANUARYFEBRUARY

MARCH

APRIL

MAY

JUNE

JULY

AUGUSTSEPTEMBER

MONTHS

Fig.3. Behavior of the Amazon river in the area of landslides.

459

OCTOBERNOVEMBER

DECEMBER

4. SLOPE FAILURE MECHANISM
The statistical analysis of Amazon River movements clearly
established that the landslides have occurred during the stageof water level decrease in the river. This is completely
different from other places around the world where rains atriver high water levels season creates landslides. We consider
that as a phenomenon of rapid drawdown that affects the bank,because of the water level decreases to an average of 12
meters in a very short time [3].
This rapid drawdown is interpreted as a process that increases
the undrained deformation of the saturated zone in theaffected banks. In other words, the reaction of the stability ofthe banks to the rapid movement when the water level
decreases is similar to the response occurred in an open cut inwhich is produced a forced alleviation, due to material that
previously was offered as lateral support and that was
suddenly removed. In this case, as a consequence of theimbalance produced by the rapid drawdown of the river, there
is water that remains within the porous structure of the soil,
since its level does not decrease to the same speed that thewater level (Fig. 4) This phenomenon causes an increase in
the weight of the bank body, as in the pore pressure with thesoil. This effect reduce the shearing strength of the soil, which,
together with the effects of the river, causes the landslides (ifit has not been possible to evacuate the water tricked within
the soil of the bank).

Fig, 4. Rapid drawdown when water level decrease
5. LANDSLIDES CONTROL MEASURESThe system of installed deep drainage is efficient and it has
generated an adequate drainage during the critical stage ofdrawdown of the Amazon river in 1996, 1997 and 1998. In
the better behavior area we put 31 horizontal drains of
30.meters of length, spaced each 3 meters with a slope of 3oand diameter of 4". In the adjacent section we installed wellswith radial drains that arrived to lengths understood between
15 to 25 meters
The measures analyzed indicate a small displacement in
direction to the Amazon river in the stage of water level
decrease, and backward displacement when the water levelrose. The comparison of the results obtained demonstrate thatthe movements registered before have reduced considerably,

460

probably due to the effective operation of the deep drainage
system, and the additional effects produced by the
sedimentation that originated due to movement of theriverbed of the Amazon river.
The results of the final piezometrics measures indicate that, asa rule, the dissipation of the pore pressures in almost all cases
has been effected in correspondence with the decrease andincrease of the water level. So, we found a good behavior in
the drainage system installed in the critical zones. The
piezometers that were installed in the zone of the lastlandslide from the beginning of their readings showed
irregularities with respect to the dissipation of the accrued
pore pressures after of the decrease of the river. It must be
noted that in the location zone of these instruments was notpracticed any deep drainage system or treatment for
maintenance [5].
6. CONCLUSIONS
The deep drainage system, installed in the studied area (bymeans of wells with radial drains as well as by horizontal
drains) has contributed effectively in the stabilization of thesebanks, and the analysis of all the measures taken during theseveral months of work with the instruments, prove that there
is a substantial improvement in the stability conditions of theplatforms included in the study, conditions that can improvein the future due to more sedimentation that presumably could
be produced in the place by effect of the change of the
Amazons riverbed.
The results shown in this paper provide a global vision of thestability problems of soils in the Peruvian wet tropic,
generated by the changing morphology of the rivers thatoriginate important risk situations in some cases, andincreasingly growing stability in others that permits toestablish the development of new behavior standards for theriverbanks of the Peruvian Amazon that in the future can bepredictable with certain aproximation considering theirevolution in the geological time of hundreds of years, since
now in certain areas it has already passed the danger, andmaybe within 100 or more years, the problem return to bepresent and the safety factors of the banks decrease gradually
until to become unstable and to produce large landslides as
they occurred in sites and dates of study, considering finallythat the Peruvian Amazon is located in a region of a very
singular world in light of their geotechnical occurrences andof climate that create very difficult wet tropical soils to
predict and handle in the construction of the earth works.
7. REFERENCES[1] Carrillo-Gil, A. “Stability Problems in Iquitos, Peru”.
Proc. VII Pan-American Conference on Soil Mechanics
and Foundation Engineering, Vancouver, Canada, 1983.
[2] Carrillo-Gil, A. , Carrillo, E. , Cardenas, J. , “Propertiesof the Peruvian tropical soils”. X Pan-American
Conference on Soil Mechanics and Foundation
Engineering, Guadalajara, Mexico, 1995.

[3] Carrillo-Gil, A. , Dominguez, E. “Failures in Amazon
riverbanks, Iquitos, Peru”. Seventh International
Symposium on Landslides, Trondheim, Norway, 1996.
[4] Carrillo-Gil, A. “Peculiarities of tropical saprolitic soilsof Peru”. XIV International Conference on Soil
Mechanics and Foundation Engineering, Hamburg,
Germany, 1997.
[5] Carrillo-Gil, A. “Geotechnology of the Peruvian tropicalsoils”, (in progress) Ricardo Palma University, Lima,
Peru, 2011.

461

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Back-analysis of Frictional Jacking Forces Based on
Shear Box Testing of Excavated Spoils
Choo Chung Siung1 and Dominic Ong E. L.2
Swinburne University of Technology (Sarawak Campus), Malaysia

1,2

ABSTRACTTrenchless technology has become the preferred technique for installation of buried infrastructure due to advantages
such as minimised ground surface disturbances as well as reduced social disruptions. These benefits have led to anincreased demand for trenchless technologies such as the pipe-jacking method, with its application well-suited forurban environments. The construction of trunk sewers in Kuching, Sarawak has been carried out using the pipe-jacking
method. Critical to this micro-tunnelling technique were the associated frictional forces anticipated during the jacking
process. Excessive jacking forces could damage installed pipes. The Authors have studied the use of a few
well-established empirical equations to validate frictional resistance during jacking supplemented by laboratory direct
shear box tests carried out on spoils obtained from pipe-jacking drives. The tested spoils were by-products of the slurrytransport system for spoil removal at the tunnel face. Shear box test results have been attempted for use to predict thepipe-jacking forces based on the well-established empirical equations. The predicted jacking forces for this study show
fair agreement when benchmarked against actual, measured forces. However, it is the intention of the Authors to carryout more tests to further verify the reliability of this method for a wider variety of local rock types, especially ingeologically ‘young’ weathered rock where competent rock cores are usually difficult to obtain.
Keywords: Pipe-jacking, direct shear, back-analysis
1. INTRODUCTION
There has been a growing need for trenchless technology forthe construction of various pipelines. This trend has been
evidenced by its implementation in the installation of trunk
sewers in Kuching, Sarawak. Sewer lines comprising of
concrete pipes were installed using micro-tunnelling by
pipe-jacking. This method was elected over conventional
open trench methods due to the minimised socio-economic
disruption and pollution. However, the technicalities of suchtunnelling works required specialist expertise. During thejacking process, various components were necessary for theworks to proceed. The capacity of the microtunnel boringmachine (MTBM), the application of lubrication, the capacityof main jacks, the use of intermediate jacking stations, and thedesign of jacking pipes are constituents of the synchronous
system for successful pipe-jacking. A key factor affecting theabove components is the frictional jacking load.
2. BACKGROUND
Studies have been carried out to investigate the behaviour ofjacking loads during the pipe-jacking process. Pioneeringstudies on pipe-jacking were carried out by researchers at
Oxford University, U.K. Norris & Milligan [1] detailed drives
entailing instrumented concrete jacked pipes were carried out.
Instrumented pipes were outfitted with pore pressure probes,
contact stress transducers, extensometers, and othermonitoring equipment. Data on variations of normal stresses
and pore pressures on the instrumented pipes provided
valuable early insight into the effects of lubrication on jacking

stresses [2, 3]. Pellet-Beaucour & Kastner [4] studied theinfluence of various parameters on the build-up of frictional
forces for drives in varying soil conditions across France. The
authors proposed two frictional force models in their analyses.
Staheli [5] studied the behaviour of pipe-soil interface friction
by conducting laboratory shear box tests. Sand specimens
were applied normally on pipe crown segments of varying
roughness. A similar test was carried out by Shou et al. [6],with the addition of lubrication. Interface frictionalcoefficients were derived from the shear box tests andverified using three-dimensional finite element modelling.
The experiences and results reported by these researchers
have been carefully reviewed and understood, so as to ensurethat the methodologies reported in this paper are within
acceptable conditions and practice.
3. SHEAR BOX TESTING OF EXCAVATED SPOILS
2.1 Direct shear test and geology
Due to the weathered nature of the ‘young’ sedimentary andmetamorphic rock of the Tuang Formation in the Kuching
City area, intact and competent rock cores (≥100mm inlength) are usually difficult to obtain during Soil
Investigation works, thus making the strength parameters (c’and φ’) for these weathered rocks rather elusive. Rock Quality
Designation (RQD) may not be the best indication of rockstrength as the values are usually zero or close to zero due toits friableness. However, this observation does notnecessarily represent the in-situ strength of the rock mass as awhole. Hence, performing direct shear tests where excavation
spoils are reconstituted with appropriate confining stress462

during testing may be a worthwhile and feasible method.
Direct shear test was conducted using the GeoComp
ShearTrac-II direct shear system (Figure 1). The ShearTrac II
system enables full automation of shear box testing.
Automation of the test ensures repeatability of testing regime.

2.2 Collection of excavated spoils
Excavated spoils were collected from the slurry separation
plant (also known as desander). The excavated spoils weretransported from the MTBM face through suspension in apressurized slurry transport system and discharged in thedesander. The desander recycles the slurry through filtrationand segregates the spoils from the slurry. The slurry is thenpumped back through the slurry supply line, back to the TBM
face. Figure 2 shows approximate chainages at which
excavated spoils were collected for testing.
2.3 Shearbox testing and results

Figure 1 ShearTrac II direct shear system
Control of the ShearTrac II system is achieved through theuse of the SHEAR software. Users are given control over the
shear box testing regime by inputting the desiredconsolidation stresses and shear strain rates. Results fromtests are generated regularly in terms of horizontal andvertical displacements, as well as shear stresses along theshear band. The ability of the user to obtain results at ahigh-resolution allows for close scrutiny of tests. The
ShearTrac II system is also able to maintain the consolidationand confining pressures throughout the test, from theconsolidation stage through to the shearing stage.

Test samples were prepared according to AS1289.6.2.2-1998[7]. Scalping of samples was carried out according to theallowances stated in the above-mentioned standard. The test
samples consisted of grains passing the 2.36mm sieve, andretained by the 75μm sieve. Such scalping method has alreadybeen carried out by Bagherzadeh-Khalkhali & Mirghasemi[8]. Scalping was deemed necessary as preliminary shear box
testing showed large fluctuations, particularly beyond thepeak stress threshold. The consolidated drained shear box
tests were conducted with samples placed in the reservoir
filled with distilled water. Shearing was initially carried out atthree confining pressures σ 1 , i.e. 170kPa, 180kPa, and
190kPa. These were approximate in-situ stresses acting on thepipe. Two further sets of tests were conducted at confiningpressures of 100kPa and 250kPa, for better data distribution
when measuring the respective strength characteristics.
Table 1 Results from shearbox testing on excavated spoils
SamplenumberSample 1

Tunnel progress points at
which spoils were collected

Sample 1

Metagraywacke

Receiving shaft

9.5m

Jacking shaft

Silty clay

Sample 2

Sample 2

38m
78m

Sandstone

103m*Not to scale
Figure 2 Estimated locations where Sample 1(metagraywacke) and Sample 2 (sandstone) werecollected as excavation spoils from slurry

σ1
(kPa)
100
170
180
190
250
100
170
180
190
250

Peak strengthcharacteristics
τ peakϕ peakc peak
(kPa)
(°)
(kPa)
117
179
171
36
49
186
227
183
282
282
49
78
282
360

Residual strengthcharacteristics
τ res
ϕ res
c res
(kPa)
(°)
(kPa)
96
128
128
28
37
134
178
112
201
203
47
4
204
275

Table 1 above tabulates the results from shear box tests on the
spoils. Generally, higher confining pressures, σ 1 resulted inlarger shear strengths, τ peak and τ res . Strength characteristics
were interpreted according to the Mohr-Coulomb failure
criterion. Sample 2 showed higher strength characteristics
than Sample 1 at both peak and residual strains. This is
attributed to the particle shapes of the respective samples,with Sample 2 exhibiting generally angular particles ascompared to the rounded particles of Sample 1 (Figure 3). As
expected, peak strength parameters were generally higher
than the residual counterparts for both samples. This isparticularly evident for Sample 1. For Sample 2, the decreasein strength from peak to residual states was significantly due
to a loss of cohesion. There was minimal observed decrease infriction angle of Sample 2. The residual strength of the463

samples is thought to mimic the excavation process better as
the in-situ rock mass is expected to be crushed till failure by
the roller cutter bits found at the face of the MTBM. Besides,
the continuous jacking process involved in the installation ofpipes is also analogous of a shearing process in large–strain
condition. The test results obtained were subsequently
utilized for the prediction of frictional jacking forces.

Table 2 Summary of various parameters for use withfrictional model by Pellet-Beaucour & Kastner [4]
b

δ

K


 π φ De 1 + 2 tan −  
 4 2 ϕ

1

Table 3 Friction coefficient values for various friction
states from Stein (as summarized in Ref. [4])Concrete on gravelor sandConcrete on clay
Asbestos orcentrifugedconcrete on gravelor sandAsbestos orcentrifugedconcrete on clay

Figure 3 (left) Sample 1 (spoils of metagraywacke); (right)
Sample 2 (spoils ofsandstone) used for direct shear tests
3. ANALYSIS OF FRICTIONAL JACKING FORCES
In the pipe-jacking process, jacking forces comprise of soilpressures acting on the face of the MTBM and normal
stresses accumulating along the outer peripheral surface ofthe pipeline. The frictional component nominally forms the
major portion of the jacking loads as the pipe-soil contact area
increases with jacking progress.
3.1 Frictional jacking force equation
Frictional forces have been the subject of much research. The
work carried out by Pellet-Beaucour & Kastner [4] employed
the model shown in Eq.(1), which is expressed as a functionof vertical soil stress on the pipe crown, σ EV as in Eq. (2).

γDe 
γD 

 + K  σ EV + e  (1) σ EV +
2 
2 
2 

h
− 2 K tan δ b(γ − 2bC ) 1 − e b(2)
σ EV =

2 K tan δ F = µLDeπ 

L is the length of the pipe string, D e is the outer diameter ofpipe, γ is the soil unit weight of ground cover, h is the heightof cover at the pipe crown, K is the coefficient of lateral soilpressure, δ is the angle of wall friction in the shear plane, b is
the influencing width of soil above the pipe. μ is thecoefficient of soil-pipe friction, expressed as(3)µ = tan δPellet-Beaucour & Kastner [4] suggested upper limit values δ= ϕ, and lower limit values of δ = ϕ⁄2 or ϕ⁄3 for soil-structureinteraction calculations. They also reported values for b, δand K by various organizations (Table 2).
Staheli [5] and Pellet-Beaucour & Kastner [4] further
reported the use of some values of friction coefficient, μ.
These values were obtained from Stein et al. [9].

Staticfriction

Slidingfriction

0.5 – 0.6

0.3 – 0.4

0.3 – 0.4

0.2 – 0.3

0.3 – 0.4

0.2 – 0.3

0.2 – 0.3

0.1 – 0.2

Fluid friction(lubrication)

0.1 – 0.3

3.2 Case study for back-analysis
A case study was considered for a drive negotiating throughthe geology of the Kuching inner city area. The pipeline lies
embedded in rock, with the crown at a level 9.5m below
ground surface. The 103m alignment of 1.2m diameter
concrete pipes traversed metagraywacke (metamorphic) andsandstone synonymous with the Tuang formation. As
reported by Tan [10], the metagraywacke encountered in the
Tuang formation is characterised as weathered, withoccurrences of quartz veins. This was consistent with findings
from soil investigation works. Metagraywacke in the Tuang
formation is often found interbedded with moderately
weathered phyllite.
The tunnel was excavated using an Iseki Unclemole SuperTCS 1200. The closed-face MTBM had an outer diameter of
1.472m, creating a theoretical overcut annulus of 19mm
around jacking pipes of outer diameter 1.434m. As mentioned
earlier, the MBTM used the slurry transport system for
maintaining a stable tunnel face, as well as for thetransportation of excavated spoils. A shield embodied themachinery in the MTBM, with a succeeding sleeve attached.
The cutter face was outfitted with four roller cutters arrangedon the face, and three roller cutters at the peripheral forformation of the overcut.
Site data of measured jacking loads (Figure 4) from the earlier
described site was used for benchmarking against predicted
frictional forces. Site data could be generalised under two
drive sections. The jacking forces within the initial drive
portion (Section 1 – 0 to 60m) accumulated within a lower
bound of 7kN/m and an upper bound of 12kN/m. The latter
drive portion (Section 2 – 60m to 108m) exhibited gains infrictional loads of between 71kN/m (upper bound) and
41kN/m (lower bound). The positions where the samples
were collected from are entailed in Figure 4 below, including
the respective shear box test results.

464

Sample 1
ϕpeak = 36°
cpeak = 49kPa
ϕres = 28°cres = 37kPa

Sample 2
ϕpeak = 49°
cpeak = 78kPa
ϕres = 47°
cres = 4kPa

Figure 4 Back-analysed upper and lower bound jacking
forces with respect to measured values including effect of
stoppages in jacking in Section 2 only
Using Eqs. (1), (2) & (3) together with the residual strength
properties of the excavated spoils, the required frictionalcoefficients for back-analysis were determined and presented
in Figure 4.
The results were based on the Terzaghi set of parameters from
Table 2. For the initial portion of the drive, the back-analysed
frictional coefficients were 0.19 (lower-bound) and 0.21
(upper-bound), while those for the latter portion of the drive
were 0.30 (lower-bound) and 0.42 (upper-bound). Withreference to Table 3, the lower-bound coefficients of friction
lie within the range of values reported for lubricated drives.
The upper-bound frictional coefficient lies beyond the rangeof acceptable values for lubricated drives.

Figure 5 Back-analysed upper and lower bound jacking
forces with respect to measured values, but with effect of
stoppages excluded from Section 2 only
resistance and effects of lubrication are usually mitigated at
this initial stage of work. This is confirmed by the
back-analysed coefficients, which are comparable to those forunlubricated sliding friction for concrete on gravel or sand as
per Table 3.

3.3 Effect of stoppage on jacking forces
In Section 2 of the drive, the disparity between upper andlower bounds for the coefficients could be explained by the
stoppages in jacking works. Jacking forces in this latter
section have been segregated as shown in Figure 5, thus
Figure 5 shows jacking forces associated with only on-goingjacking works. Frictional coefficients for the lubricatedSection 2 are seen to have reduced to 0.32 (lower-bound) and
0.34 (upper-bound). These coefficient values were slightly
above the upper limit of the corresponding values reported inTable 3 (under lubrication component). Figure 6 shows
jacking forces, with forces attributed to stoppages presented
for Section 2 of the drive. Frictional coefficients for Section 2
are 0.36 (lower-bound) and 0.42 (upper-bound), which are
generally higher as re-start of jacking works after periods ofstoppages are known to induce greater static frictional

Figure 6 Back-analysed upper and lower bound jacking
forces with respect to measured values, with only effect of
stoppages in Section 2 only

465

Concrete on
gravel or sand(sliding)

pipe-jacking process as had been described in detail earlier
and interpreted from Figures 4, 5 and 6.
Figure 8 illustrates the occurrences of extended stoppages,
where sudden increments in jacking forces have beenpreviously presented. These were not weekend stoppages as
jacking was on-going through the weekends. Such stoppages
in work could have been due to major maintenance works,
possibly from the replacement of roller cutters. During suchstoppages, lubricant injected into the overcut around the pipe
peripheral could be lost through fissures in the surrounding
geology as shown schematically in Figure 9.

Concrete on
gravel or sand(static)

Section 2 (during jacking)μ = 0.34 (upper)
Section 2 (stoppages)
μ = 0.42 (upper)

Section 2 (during jacking)μ = 0.32 (lower)

Section 2 (stoppages)
μ = 0.36 (lower)

Section 1 (on-going)
μ = 0.21 (upper)
Section 1 (on-going)
μ = 0.19 (lower)

0

0.1

0.2

0.3

0.4

0.5

0.6

Partial lubricationin overcut due tofissures

Loss of lubricant
into fissuresμJackingpipe

Full lubrication in
overcut

Fluid friction(lubrication)
Concrete-clay(sliding)

Concrete-clay(static)

Figure 7 'Dynamic' behaviour of coefficient of friction
during various stages of the pipe jacking process

Figure 9 Schematic idealisation showing
lubrication through fissures in a rock massloss ofThe pressure for lubricant injection would also be relieved,
mitigating any buoyancy effect that may have reduced thefrictional forces. The effect of loss of buoyancy supporting
the weight of the pipe could result in higher pipe-soil contact
area. There was an observed change of the lubricating regime
in Section 2 of the drive, particularly upon the advent of thethird major stoppage. This increase in the volume of lubricant(from 170 litres to 800 litres per metre run) was in reaction to
the increased rate of frictional loads upon transition from
Section 1 to Section 2 of the drive. However, this intensified
effort had only been initiated after several spikes in thejacking load had been accrued.
4. CONCLUSIONS

Figure 8 Development of jacking force and elapsed timein relation to stoppages and volume of lubricant injected
into tunnel overcut
Figure 7 shows the ‘dynamic’ propagation of friction
coefficients during different stages of work in theThis paper has highlighted the potential of using direct shear
testing of excavation spoils to obtain residual strength valuesof geologically ‘young’ weathered rock to predict jacking
forces. The main advantage of using reconstituted excavation
spoils as specimens is the ability to obtain rock strengthparameters which otherwise would be elusive as intact andcompetent rock cores (≥100mm in length) are usuallydifficult to obtain in such weathered and friable conditions.
One of the main factors affecting the accuracy of the
back-analysis of the lower and upper bound jacking forces isthe ‘dynamic’ or fluctuation nature of the coefficients offriction as the pipes penetrate into differing geology within asimilar drive and also due to its construction activity (whether
jacking is on-going or stopped for maintenance). The
back-analysed coefficient of friction seems to be also affectedby the fact that whether the drive is fully or partially466

lubricated. Hence, soil-structure interaction is indeed an
important consideration in any pipe-jacking work.
Jacking forces have been observed to have increased during
the re-start of jacking activities after periods of stoppages as
more thrust is required to overcome static frictional resistanceprior to the realisation of the positive benefits of lubrication.
However, it is the intention of the Authors to carry out moretests to further verify the reliability of this method for a wider
variety of local rock types, especially in geologically ‘young’
weathered rock.
5. ACKNOWLEDGMENTS
The authors would like to thank the generosity of Hock Seng
Lee Bhd and Jurutera Jasa (Sarawak) Sdn Bhd for theirsupport during this study. The authors are also thankful to theground staff of Shaft & Tunnel Sdn Bhd for their assistanceand patience during the study of the microtunnelling drive.
The work described in this paper would not have beenpossible without the efforts of Leong Hsiao Yun and Lionel
Foo Fang Ho.
6. REFERENCES[1] Norris, P. and G.W.E. Milligan, Field instrumentation
for monitoring the performance of jacked concrete pipes,
in 3rd International Symposium on Field Measurementsin Geomechanics, G. Sørum, Editor 1991, A.A.
Balkema: Oslo, Norway.
[2] Milligan, G.W.E. and M.A. Marshall. The functions andeffects of lubrication in pipe jacking. in World Tunnel
Congress. 1998. Sao Paulo, Brazil: Balkema.

[3] Milligan, G.W.E. and M.A. Marshall. The influence of
lubrication on jacking loads from six monitored pipe
jacks. in 16th International ISTT No-Dig 1998. 1998.
Lausanne, Switzerland: ISTT.
[4] Pellet-Beaucour, A.L. and R. Kastner, Experimental andanalytical study of friction forces during microtunneling
operations. Tunnelling and Underground SpaceTechnology, 2002. 17(1): p. 83-97.
[5] Staheli, K., Jacking force prediction an interface friction
approach based on pipe surface roughness, in School of
Civil and Environmental Engineering2006, Georgia
Institute of Technology: Georgia. p. 385.
[6] Shou, K., J. Yen, and M. Liu, On the frictional propertyof lubricants and its impact on jacking force andsoil–pipe interaction of pipe-jacking. Tunnelling andUnderground Space Technology, 2010. 25(4): p.
469-477.
[7] The Australian Standards, AS1289.6.2.2-1998 Soilstrength and consolidation tests - Determination of theshear strength of a soil - Direct shear test using a shear
box, 1998, Standards Australia: Australia.
[8] Bagherzadeh-Khalkhali, A. and A.A. Mirghasemi,
Numerical and experimental direct shear tests forcoarse-grained soils. Particuology, 2009. 7(1): p. 83-91.
[9] Stein, D., K. Möllers, and R. Bielecki, Microtunnelling :Installation and Renewal of Nonman-Size Supply and
Sewage Lines by the Trenchless Construction Method.
1989, Berlin, Germany: Ernst.
[10] Tan, D.N.K., Geology of the Kuching Area, West
Sarawak, Malaysia, 1993, Geological Survey of
Malaysia: Kuching.

467

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Lime Stabilization of Tropical Soft Soils
Dygku Salma Awg Ismail1, Tan Jui Siang2 and Siti Noor Linda Taib 3
Universiti Malaysia Sarawak, Malaysia

ABSTRACTSoft soil, naturally contains high percentage of moisture content, is having low values of compression strength andCalifornia Bearing ratio (CBR). Construction of any infrastructures over the soft soils unsurprisingly causes the soil toundergo excessive settlements and possible bearing capacity failures. Therefore, chemical stabilization has been practiced
by the practitioners whereby the chemicals such as cement and lime were used in construction. In this study, lime
stabilization method was opted to improve the geotechnical properties of the soft soils. The initial consumption lime tests
were done at the first stage in order to determine a minimum amount of lime used for the stabilization. In addition, theunconfined compression strength (UCS) and California Bearing Ratio (CBR) tests were conducted to evaluate theeffectiveness of lime stabilization with different percentage of lime content. As a result, the highest UCS and CBR valueof stabilized soft soils was achieved at 9% of lime content within 28 days curing period which indicated a value of 1514.13
kPa and 92.19% respectively.
Keywords: soft soils, curing period, lime stabilization, optimum lime content
1 CONCLUSIONSoft soils are commonly tag as ‘problematic’ due to its
typical failure characteristic of having poor resistance todeformation, highly compressible, low shear strength, lowpermeability and limited bearing capacity [1]. Construction of
any infrastructure over soft soils unsurprisingly causes thesoil to undergo excessive settlements and possible bearing
capacity failure. The common construction approach in soft
soil is excavation and substitutes the soft soil with stronger
and better quality materials. However, this effort will lead to
uneconomical design due to the requirement of purchasing
and transportation for large amount of good quality soil [2]. Insuch conditions, alteration of the weak properties of soft soil
to meet the specified engineering requirement can beachieved by adopting soil stabilization. Soil stabilization canbe defined as the permanent physical and chemical alterationof soils to enhance their physical properties. Stabilization can
increase the shear strength of soil and control the shrink-swellproperties of a soil, thus improving the load bearing capacityof a sub-grade to support pavements and foundations [3]. The
techniques of soil stabilization can be classifies into a numberof categories such as vibration, surcharge load, structural
reinforcement improvement by structural fill, admixtures,
grouting and etc. [4].
2 RESEARCH METHODOLOGY
Hydrated lime Ca(OH) 2 was used as the chemical additive
for stabilization of this study. The quality and suitability of
hydrated lime used has to be ensured before the stabilization
process. Table 1 shows the soil properties and Table 2

illustrates the chemical properties of the hydrated lime used inthis study. A light storm given was reported to have occurreda day prior to sampling which influenced the moisture content
as indicated in Table 1.
Several physical properties testing were conducted on soil
and hydrated lime in order to determine the suitability ofthese admixtures. It was also important to conduct these tests
as a control set to compare the soil properties before and after
soil stabilization. After verification of soil properties andsuitability of the soil and lime admixtures, estimated lime
content was added to the soil for stabilization.
In this study, lime content of 7, 8, 9 and 10 percent by soil
weight were used. Unconfined Compressive Strength test
(UCS) and California Bearing Ratio test (CBR) were carriedout on the 28 days air-cured samples to evaluate theeffectiveness of lime stabilization with different
concentration of lime solution. The Atterberg limit test andmoisture content determination were then conducted as toinvestigate the effect of lime stabilization on plasticity
behavior of the soil.
2.1 Initial Consumption Lime
A series of initial consumption lime tests (ICL) werecarried out in which to determine the ICL value for theselected soil. The ICL value is indicating a minimum amountof lime required for soil modification in the early stage.
Therefore, the lime used in the stabilization process is greaterthan the ICL value in order to establish soil stabilization. The
ICL value was determined based on average pH of 12.4 inorder to sustain the strength producing lime-soil pozzolanic468

reactions. As a result, about 6% of lime was obtained from the
test.
Table 1 Properties of soil
Physical properties
Natural Moisture Content (%) before
32.78
light storm
53.12
Natural Moisture Content (%) after
light storm
Specific Gravity (G s )2.64
Liquid Limit (%)
61.50
Plastic Limit (%)
30.15
Plasticity Index (%)
31.35
Unconfined Compressive Strength
167.80(UCS) test (kPa)
California Bearing ratio (CBR) (%)
10.08
Figure 1 Standard Proctor Compaction Curve

Particle size distribution
Sand
Silt
Clay
Clay activity (A c )

69.50
16.70
13.80
2.27

Soil Classification
BSCS
AASHTO

SCH
A-2-7

As in Fig. 1, the compaction curves show that further
addition of lime to the soil can decrease the dry density and at
the same time increase the moisture content of the mixture.
Moreover, the compaction curves move closer towards thezero air void line as the lime content is increased.
3 RESULTS

Table 2 Chemical properties of hydrated lime
Available Lime Content
Ca(OH) 2
85.71
CaO
64.86
pH at corrected 25 ºC

12.40

2.2 Optimum Moisture Content
Optimum moisture content determination is important as acondition to prepare a fully saturated condition for lime
stabilization. The result of standard proctor compaction test is
given in Table 4 and the compaction curves for various lime
contents are plotted in Fig. 1.
Table 4 Standard Proctor Compaction
Lime Content
Maximum Dry Optimum(% of soil mass)
Density (Mg/m3) Moisture
Content (%)
1.44
25.5
Untreated
1.35
29.2
7%
1.32
33.5
8%
1.30
33.9
9%
1.28
34.2
10%

3.1 Unconfined Compressive Strength (UCS)
Compressive strength of a soil is a significant factor toestimate the design criteria for the use as pavement andconstruction material. The result of the UCS test for the soil
samples at different lime content are tabulated and shown in
Fig. 2. The highest strength achieved is 1514.13 kPa with 9%of lime content cured for 28 days at room temperature of 27 ±2ºC. The reason for the strength development is due to the
pozzolanic reactions between lime, water, soil silica, andalumina that form various cementing-type materials. The
cementing products are calcium silicate hydrates and calcium
aluminate hydrates which are the same hydrates formed
during the hydration of Portland cement [6]. According to [7],the excessive lime can be detrimental in soil stabilization if
the available silica in the soil is consumed. Fig. 2 shows thestrength of soil-lime admixtures are considerably increased
according to the increment of lime. Nevertheless, thecompressive strength of the treated soil shows no further
increase after optimum lime content was achieved.

469

of the soil sample, which results in a significant reduction in
the plasticity index. This reduction in plasticity index means amarked increase in workability. The effect of lime on liquid
limit is much less marked compared to the plastic limit. As
indicated by Fig. 4, addition of 7 % of lime content increases
18.23 % of plastic limit while the liquid limit only decreases
by 1.91%.

Fig. 2 Unconfined Compressive Strength Test

3.2 California Bearing Ratio (CBR)
California Bearing Ratio (CBR) value is used as an index ofsoil strength and bearing capacity. It is a familiar indicator
test used to evaluate the strength of soils which broadly
applied in the design of the base and the sub-base material forpavement [8]. CBR test without soaking are conducted tocharacterize the strength and bearing capacity of the soil
samples with different lime percentage compacted at
optimum moisture content. Fig. 3 gives the summarized resultof CBR value for the untreated and lime treated soil. The
results of unsoaked CBR values for lime treated soil indicatedthat the highest CBR of 92.19 % is observed at 9% lime
stabilization. The CBR values are increasing as the lime
percentage is increasing up to 9% and thereafter the CBRvalues started to reduce. Thus, lime stabilization isrecommended at 9 % lime content.

Fig.4 Influence of the Addition of Lime on Atterberg Limit

Fig. 5 presents the reduction of water moisture content of theuntreated and lime treated soil after 28 days air-cured at room
temperature of 27 ± 2ºC. From the result, the moisture content
of the soil reduced 14 % at the lime content of 10 %. It showsthat drying occurs only through the chemical changes in thesoil that reduce the soil moisture holding capacity andincrease its stability as for hydrated lime [10].

Fig. 3 California Bearing Ratio

The improvement of a cohesive material using hydrated
lime changes the clay properties in plasticity index [9].
Atterberg limit test are conducted for lime treated soil as tocompare the variation and differences with the untreated soil.
Fig. 4 summarized the effect of the addition of lime on theplasticity of the soil sample. From the results, it is clearlyshown that the addition of lime to the soil increases the plastic
limit however decreases the liquid limit and linear shrinkage
470

Fig. 5 Moisture Reduction of Lime Treated Soil

4 CONCLUSION
In this study, it was found that the 28 days air-cured lime-soil
admixtures contributed the highest strength at 1514.13 kPa
with 9% of lime content. In addition, the unsoaked CBRvalues for lime treated soil also indicated the highest CBR of
92.19 % at the same value of lime content. Besides, there isno further increase of strength after the stabilization reaching
9% of lime content. Therefore, the optimum concentration oflime used in this study is 9% of lime by weight of the soil.
5 ACKNOWLEDGMENT
The authors acknowledge Universiti Malaysia Sarawak forthe support towards this project through UNIMAS GRANT
02 (76) /581/2006/(14).
6 REFERENCES[1] Gan, S., & Tan, C. (2003). Some Construction Experiences on Soft
Soil Using Light Weight Materials. 2nd International Conference on
Advances in Soft Soil Engineering and Technology. Putrajaya,
Malaysia.
[2] Said J., & Taib, S. (2009). Peat Stabilization with Carbide Lime.
UNIMAS E-jornal of Civil Engineering Vol. 1.
[3] ARRA, A. (2008). Soil Stabilization. Retrieved October 2011, fromwww.arra.org: http://www.arra.org/index.php?option=com_content&
view=article&id=36:soil-stabilization&catid=20:industry-segments&
Itemid=81.
[4] Kazemian, S., & Huat, B. (2010). Assessment of Stabilization
Methods for Soft Soils by Admixtures. International Conference on
Scientific & Social Science Research (CSSR 2010). Kuala Lumpur,
Malaysia.
[5] McNally, G. (1998). Soil and Rock Construction Materials. London:
Routledge.
[6] Terrel, R., Epps, J., Barenberg, E., Mitchell, J., & Thompson, M.
(1979). Soil Stabilization in Pavement Structures: A User's Manual,
Volumes I and II, FHWA.
[7] Impe, V. (1989). Soil Improvement Techniques and Their Evolution.
CRC Press, Taylor and Francis.
[8] Nicholson P., K. V. (1994). Lime and fly ash admixture improvementof tropical Hawaiian soils. Washington DC: Transportation Research
Record.
[9] Transport, D. o. (2007). Treayment of Fill and Capping Materials
Using Either Lime or Cement or Both.
[10] NLA, National Lime Association. (2004). Lime-Treated Soil
Construction Manual: Lime Stabilization & Lime Modification.

471

Increase Of Piles Bundle Capacity By Making
A Base Joining Them

Jan Jaremski
Rzeszow University of Technology, Rzeszów , Poland, jjaremsk@prz.edu.pl

KEYWORDS: joining pile, rock eluvium, swelling soils.

ABSTRACT: The paper contains a new solution of the complex pile foundation including
the base, made in the ground, and the joining piles the base with the girt. This kind of piles canbe used for foundation under special geological conditions (proglacial stream valleys) for foundation
reinforcement of the existing buildings and new foundations. The proposed solution may be appliedin the swelling soils. . In the work possibilities of execution of joining piles in different soils likefine sands, silts, clays, clay shale, sandstones, which can be foundation for the pales, have been
considered. Application of the joining pile on swelling soils.
of piles for improvement of soil capacityof theexisting
foundations
locatedon the grounds of low load capacity andon swelling ground. In many cases, direct
capacity of the considered pile canbe compared with high-diameter piles
because of theirbase surfaces.
The considered solution is very cheap
and its realization does not require heavyequipment at the building site. It also
eliminates a widely applied „forest”
of prefabricated piles joined with a plate
requiring much more reinforcement thanbeams joining the joining piles.
The joining pile is characterized
by great ability to transfer horizontal loads,
very important in bridge and industrial
building (chimneys etc.). It can be also
applied as the anchor pile in small buildings.
The considered piles are used for increaseof capacity of the existing foundationsin the caseofasuperstructureand deterioration of geotechnical parametersof the base.
The discussed way of foundation was
realized in the soils where a layer of gravelsand soils occurred below the layers non-

1 INTRODUCTION
So-called joining piles represent a new ideaof piling where the basic part of the pile(the base of the high-diameter pile) is locatedat the depth of the foundation occurrence.
Depending on the ground conditions, in many
pile structures the basic part of loading
is carried by the pile base. The joining pileis an example of such piles. It is a combined
foundation including the base madein the soil, and the post elements (a bundleof piles or micropiles) joining the base withthe girt (see Fig.1).
Loading conveyance from the building object
to the carrying ground zone takes placethrough the bundle of piles (posts) joinedat the base with the solid of a big projection
zone and volume. The solid is formedof the hardened soil-concrete generated fromthe deposited aggregate or the degraded rockwith injections cement grout. Such a pileallows to distribute loading on a large
surface; in a consequence load intensity
becomes lower. It also allows to make a baseincreasing load capacity of the bundle
472

transferring loads. In such layers the pile basecan be made with cement paste injection.
Such geological conditions usually exist near
proglacial valleys and valleys where there
are interbeddings of aggradate mud, peat, soft
plastic silts clay, dust especially sensitive
to changes of humidity occurring e.g. while
flood states

In the cases realized so far, the pile bases
were made in the gravel layer occurringat the depth of about 5 m below the siteby means of an injection of cement paste.
Such injection was performed throughthe holes in a bundle containing e.g. four
posts (prefabricated piles with the holes
inside).
There is another advantage of joining piles:
they can be used at expansive soils. Suchpiles have relatively very small lateralsurface. Moreover, it allows to transfer someforces from swelling through anchors locatedin the holes used for injections. Foundation
on traditional piles in swelling soils
transferring loadings to non-swelling layers
does not protect the structures against
the swelling effects because of swellinginfluence on lateral surfaces of the piles. The
joining pileisaverifiedsolutionfor foundation reinforcement in the existingbuildings, it can be applied for foundation
capacity increase in dangerous situations,also in the case of new solutionsof intermediatefoundations[7].
The calculations
includedpossibilitiesof deviations
inpost
elements
with unintendend eccentricity of longitudinal
force applying. For calculations parametersof the stratified base were assumed, becausesuch base usually occurs in proglacial
valleys. The base joining post elements canbe done only under the condition of existenceof suitable layer. Thus, the proposed idea canbe realized only in proglacial valleys[6].
Application of the joining piles under
different conditions has been considered.
Namely, it concerns the conditions where
there is no layer of non-cohesive soil,
and at the level of pile foundation there
are shale clays, shales, sandstones, finesands, silts, clays etc., being a base forthe element joining the pile heads. It seems
be possible to make an empty spacenecessary for generation of the pile baseby explosion of an explosive material.

Fig. 1. A scheme of foundation reinforced with abundle of piles joined by concrete base473

The formed volumes could be filledwith cement grout[8].
In the considered solution, explosive
materials are used while pile manufacturing.
Ammonium saltpetre and potassic saltpetre
are the proposed explosive materials,
comparable amounts of trinitrotoluene arealso considered. A possibility of such
solution can be proved by an example from
Shanghai, China. Ground anchors werelocated under the object of relatively large
area . That object contains two undergroundstoreys, and soft plastic swelling clays occur
at the foundation level. The object is locatedon the plate mounted by means of 1100
anchors 9 meters in length with a widenedbase. A necessary volume for the widenedbase of the anchor was obtained by meansof explosion of the explosive material.
In such a way, an empty area of a small
diameter was obtained at a proper depth
in a soft rock. The area was filled withconcrete. Ammonium saltpetre and potassic
saltpetre were applied as explosive materials,
and suitable detonators were used –application of professional procedures could
eliminate accidents. Vibration and noise
generating while explosion were less thanpermissible ones. Effects of explosion andgenerated shocks were analyzed. Noise
measurements done in situ while explosions
showed from 112 to 116 decibels.
In opinion of the scientists from China,
the presented solution is a new methodof foundation anchorage at expansive soils.
The applied method of manufacturing
is reliable
and
very
good
from
the economical point of view, and it should
be disseminated – the proposed design
is safe, relatively cheap and its realization
is easy[1].
In the case of soft rocks at the levelof pile location, it is also possible to makea head of soils occurring above the eluvia
of soft rocks widely assumed as non-carrying
subsoil.
The
author analyses

joining in cohesive soils and the soft rocks.
For many pile constructions weatheringis the foundation strata and it is very complex
to learn its properties and geotechnicalparameters considering the difficultyof taking samples of undisturbed structure
and the depth which weathering occupies.
The researches , given in the early papers ,can be accepted as model onesin the estimation of geotechnical parametersof weathering which is very often
the foundation strata for pile constructions[3,4,]. The obtained results of the researchesof the marl eluvium are ones of the standard
laboratory tests of debris with disturbed
structure, made successively , and alsoof model testing made on the samples
of weathering with the smallest fractionsof simulated variations of moisture -content
approximatedtothoseoccurringin the deposit.
The
paperspresented
the results of non-typical investigations“in situ ‘’ to with the role of a verifierof numericalvaluesofgeotechnicalcharacteristics has been attributed [5,10,].
They allow to determine the optimum
bearing capacity of piles which brings about
the possibility of the reduction of their
number and greater safety of designed
constructions. The determination of piles
bearing capacity is based on test loads [2,9].
2 THE COMPLEX PILE
MANUFACTIURING FOR NEW
OBJECTSIn case of many pile solutions (high-diameter
piles) the fundamental part of loadings
is transferred by the pile base. The basemanufacturing requires high-diameter boreholes made by means of heavy equipment.
Such piles are rather expensive. In suchcases, the joining pile can be an alternate
solution. In case of such a pile, at least threeprefabricated
piles
are
introduced
in a traditional way (by means of a vibrating
hammer or a pile driver) at the depthof occurrence of the water bearing layer

the possibilities of utilization of earlier
investigations of application of the pales474

3 APPLICATION OF THE PILE ON
SWELLING SOILS

containing coarse sands and gravelon the surface of min. 0.5m2. Next, injection
is performed by the axial holes in piles,
at least 20 in diameter; where the binding
medium (for example cement grout)is introduced under pressure 3 – 150 bars.
While injection into the first hole,
the moment of occurrence of cement groutin the hole of the alternate pile is determined,
and injection pressure treated as preliminarypressure is registered. .
The pile base formed after injection
and cement grout binding, joined by piles
with the girt (the plate the joining pile heads)
is a complex pile foundation. For example,
the joining pile manufactured by introductionof four prefabricated piles of dimensions
80 x 80 and the spacing about 80 cm forms
after injection the base joining pile headsof a diameter comparable with the highdiameter pile 180 cm in diameter.

Usability of the pilejoining
on the expansive soils is its important
advantage. Conventional piles locatedin swelling soils transfer loadings to nonswelling layers, but they do not protect
the structure against effects of swelling
because of action of forces from swellingon lateral surfaces of piles. The joining pile
has a relatively small lateral surface.
Moreover, it allows to transfer some forcesfrom swelling through anchors locatedin the holes for injections, diameters of which
can be enlarged. This solution was
successfully applied in two constructions.
Swelling soils are present on many areasof the hot and temperate zones, also in many
enclaves of soils and rock eluvia containing
mineralsbelongingtothegroupsof montmorillonite, illite etc. Identificationof swelling
abilityandmethods
of its investigations
weredescribed
in [11,13]. Losses caused by soil swellingin building industry are comparable to losses
caused by environment.

In the case of swelling soils, the systemof piles or micropiles with the basejoining their heads plays the roleof the anchoring pile. After the injection,
the anchors made of steel ribbed bars
are introduced
throughtheholes
at the depth
providingaproperanchorage. The fronts of the piles
are bonded with the foundation masonryor the structural post by the steel girts.
In the case of unfavourable soil
and water conditions above the base,the micropiles are made by forcing the pipe
into the base. At the pipe front there is a taperwith a hole for injection. When a proper
depth is obtained, injections of the hardening
medium are performed into the base, next
the pipe with the taper is removed. Concrete
mix is introduced into the existing hole,and the micropiles are formed.

Fig 2. The computer simulation of the joining
pale

475

Under the foundation of the consideredbuilding, the heads were made – they joined
the pile bases made by injection in the gravel
layer occurred at the depth about 5 m.
Injection of cement grout was doneby the holes occurring in the bundle of four
posts (prefabricated interval piles φ150 mm
and with the internal holes φ42 mm
and length
140 cm). .The prefabricatedbundle of piles with the base was made underparticularcores(hiddenposts)
in the foundation masonry every 3.0 m underposts of reinforced concrete frames. Such ahigh-diameter pile under each of 54 postsof the reinforced concrete frames allows tojoin a bundle of 4 posts with the cores bymeans of a steel girt (2 piles at each side).
The girt is an angle bar reinforced in threepoints and joined with the post by means of 4bolts φ50mm. .
Since swelling properties of the soils
occurring along the posts (prefabricatedpiles) are variable, the piles play
an additional role of the anchoring piles. .
After injections, the bars made of ribbed
steel, 20-40 mm in diameter, were introduced
into the holes inside the pile to the length
greater than 40 diameters. Injection
was performed under pressure depending
on density of the layer of coarse sands
and gravel occurring in the subbase.
The joining pile should contain at least
4 prefabricated piles allowing to makea symmetric injectedpart of the pile.
The prefabricated piles are introduced
at the establisheddepth.
Assembly
was realized according to requirementof the adapted building. The prefabricatedpiles are interval piles of φ150 mm –it is the maximum diameter of the drill
applied for hole drilling. The ready holes
were checked at the building site.
Introduction of the prefabricated element
is preceded by making a hole φ150mm
in the soil by meansof the tapered point chisel (see Fig.1. 3).

4 JOINING PILES MANUFACTURING
WHILE
INCREASE
OF
LOADCAPACITY OF THE
EXISTING
FOUNDATIONS
In future it will be possible to apply
the considered piles in modernized objects
made of brick. It is proposed to make thejoining
piles with the simultaneous
introduction of the beams made of steel
profiles into the masonry. The beams could
transfer loadings from the masonry to theposts-piles. If a frogged storey
must
be made, the piles could be built over. How
to make such piles when load capacityof the existing foundations is increased. Hereis an example. As an example cracks
occurred on the walls and floor cellars at the
final phase of the major overhaul of thebuilding. Cracks occurred on the walls at the
final phase of the adaptation repair of thebuilding. Also non-uniform deformations
caused by ground settlement under a part ofthe building, and the ground swelling wereregistered. It is caused by vertical cracks on
the walls in the cellar in eighteen reinforced
concrete cores. The supporting structure of
the higher storeys included reinforced
concrete frames, combined as a monolith inreinforced concrete finials of the ceilings ofchannel plates. After 25 years later ofoperating, the internal walls were damaged.
During adaptation works the walls were
replaced by new ones. The substrate below
the building should be consolidated after solong time, and according to the geodetic
measurements results a part of the object was
settling, and another part was period rising.
Particular layers of the substrate were tested
very precisely. All the tests proved
a complicated structure of the ground below
the building. The soil cooperatingwith the foundations below all the houseis and it will remain the active substratewith variable
geotechnicalparameters
because of swelling, mineralization oforganic parts of the soils occurring above thevariable level of the underground water.
476

and in a consequence load intensity canbe reduced. This method is much more
efficient than other well-known andwidely applied methods of increase of
the
soil
load
capacity.
The connection butt-end micropiles it
was how on Figure.

Fig 3. Diagram of the micropile introduction
In this order, the assembly girt is mounted
on the foundation masonry at the height
dependent
ontheservo-motor
and the element length. Next, the point chisel
was forced into the hole.

5 CONCLUSIONS

The reinforced concrete post elements
were introduced into the hole by meansof the same hydraulic servo-motor. The
main hole is made to the required depth
by means of joined sections of thickwalled tubes ~φ100 - 140 mm, joinedwith the tapered point chisel whichallows to make a preliminary hole where
the pile is introduced. When the taper
is introduced to the required depth,
the point chisel is removed by meansof the hydraulic servo-motor. According
to measurement of pressure, the forceof the point chisel introduction was about
80kN.

1. The planned works lead to the complex
analysis of the presented problems and theyshould insist in tests in situ, model tests,
numerical
calculations.
Determinationof the injected zone of ground concrete isalso provided. Determination of injection
parameters in fluvioglacial soils is also a very
important problem.
2. Application of the joining piles seems
to be advantageousascomparedwith foundation of new objects on the highdiameter piles. Such piles are very expensive,
and their manufacturing requires heavy
boring rigs and much time, also applicationof special equipment. .
3. Application of the joining piles does
not require special equipment for earth
works, especially in the case of increaseof load capacity of the existing foundations
at the soils of low capacity and swelling soils.
4. A possibility of application of the joining pileat swelling soils is its very important
advantage.
Traditionalpiles
appliedat swelling soils can transfer loadings to nonswelling soils, but they do not protect
the structure against the swelling effects
since swelling acts on the lateral surfaces ofthe piles.

Application of the considered piles does
not require special equipment for earth work,
especially in the case of increase of load
capacity of the existing foundations
on the soils
of
low
load
capacity
and on the swelling soils.

The load transfer from the building object
to the carrying soil zone throughthe bundleofmicropiles
joinedat the base by a block of a big projection
surface formed by cement grout allows
to distribute the load into a large surface,477

6 REFERENS

8. Jaremski J.,(2006) Possibilities of aplicationof rzeszowski pales In cohesive soils and thesoft rocks. O możliwościach zastosowania
pali rzeszowskich w gruntach spoistych iskałach miękkich.( in polish),.
Zeszyty
NaukowePolitechniki
Białostockiej,
Budownictwo z. 28
9. Jaremski J.: Creep property of Opole

1.

Guanzhou Z. Zhiming X.( 2004):
The application of blown tip pile usedin the saturated soft soil land single stakeeconomic analysis. Proc. Int. Conf. on Soil
Nailing, Nanjing,.
2. Gwizdała K.:(2008) Fundamenty palowe.
( In polish). Wydawnictwo Naukowe PWN
SA. Warszawa.
3.
JaremskiJ.:(1991)Characteristicof weathering as foundation strata for pile
construction on the basis of the researchesof the Opole marls eluvium. Proc. 4th Int.
Conf. Piletalk International, Kuala Lumpur,.
4.
Jaremski J.: (1997) The influenceof recognition of physical and chemical
processes occurring in the weatheringin the estimate of the extreme valuesof geotechnical parameters of the Opole
marls eluvium. Proc. Int. Symp. GeotechnicalEngineering of Hard Soils-Soft Rocks,
Athens, A.A.Balkema,.
5. Jaremski J,:(2005), The influence of physicalandchemicalprocesses
occurringin the weathered
Opole
marls
on the geotechnical parameters, Publishing
House of Rzeszów University of Technology,
6. Jaremski J.:( 2000) Increase of micropiles
bundle capacity by making a base joining
them.Zwiększenie nośności wiązki mikropali
przez
wykonaniełączącej
jepodstawy.(In polish), XLVI Konferencja
Naukowa KILiW PAN i KN PZITB,
Krynica,.
7. Jaremski J., Wilk K.: (2001) Zwiększeniemożliwości wykorzystania pali rzeszowskich
do posadowień fundamentów pośrednich(In polish), VI Międzynar. Konf. Naukowa
Lwowsko- Koszycko-Rzeszowska, Lwów.

marls weathering on the base of themodel laboratory research on the researchin situ. Proc. 7th Int. Cong. on Rocks
Mech. Aachen, A.A.Balkema, 1991.

10. Jaremski J.:(1994) Influence of physical andchemical processes occurring in eluvium of
the Opole marls and their influence on thegeotechnical parameters, Proceedings of the
7th International Congress IAEG, Lisboa ,A.A Balkema,
11. Jaremski J.:(1995) Proposal for a method
determining value of the maximum swellingof soil by example of the Opole marls
eluvium and illite clays. Proc. of 8th Int.
Cong. on Rock Mechanics, Tokyo,
A.A.Balkema.
12. Jaremski J.:(2004) The influence of changes
of weathering water content on the formationof landslides in the Carpathian Flysch. Proc.
of 9th Australia New Zealand Conference onGeomechanics, Auckland – February 2004,
New Zealand Geotechnical Society Inc.
13. Jaremski J.:(2010), Influence of moisturechanges on physico - chemical processes andgeotechnical parameters in soft rocks , Studia
Geotechnica et Mechanica, Vol. XXXII,
No. 1.

478

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The Effects of Soil-Pile Interaction on Seismic Parameters of Superstructure
Mahdy Khari1 #, Khairul Anuar Bin Kassim2, Azlan Bin Adnan3
1. Department of Geotechnical Eng., Faculty of Civil Engineering, Universiti Teknologi Malaysia
2. Department of Geotechnical Eng., Faculty of Civil Engineering, Universiti Teknologi Malaysia
3. Department of Stretcher Eng., Faculty of Civil Engineering, Universiti Teknologi Malaysia(#: Corresponding Author: Mahdy Khari)
ABSTRACT
Large superstructures such as tall buildings, important structures and bridges sometimes are founded on fluvial
and alluvial soil deposits. The pile supported structures response subjected to earthquake events strongly affected by
the dynamic behavior of the soil-pile interaction (SPI). The destruction of the pile-supported superstructures has
raised the attention to the SSI underlying concerns. In this paper, an attempt is made to evaluate the influences of thelaterally seismic excitations on the lateral response of flexible piles owing to the kinematic and the inertial
interaction in the SPI. In the present approach, the nonlinear beam on Winkler foundation (BNWF) is incorporated
into a finite element analyses program to calculate the piles response. The static p-y curves are implemented tosimulate the soil reactions within the frame of Winkler method. The material damping was included in this method.
The results obtained of the equivalent linear method of the site response analyses are utilized as the input motions.
The computed responses have been shown that this approach can be used for estimating the soil-pile interaction
effects on the system seismic behavior.
Keywords: Soil-Pile Interaction, BNWF, Seismic Parameters, p-y Curve, Earthquake

1. INTRODUCTION

Based on these observations, the several methods
have been developed to evaluate the seismicparameters of structure with considering of the soilpile interaction such as finite element method(FEM)[4-5], boundary element method (BEM)[6-7],
beam on nonlinear Winkler Foundation (BNWF)and simplified method. It is mentionable although
FEM and BEM provide the powerful tools to verify
the seismic behavior so that the soil is considered as
the continuum medium but they are not commonly
owing to the excessive computational costs used.
Among these foregoing approaches, the BNWF is
widely used in the design offices [8-9] because it isa versatile and economical method.
A BNWF analysis of the soil-pile-structureinteraction is performed in the current paper toverify the seismic behavior of the superstructureusing the general finite element analysis software.

Significant damages of superstructures supported
by deep foundations from major earthquakes (e.g.
Cairo
Earthquake,
1992;
Loma
PrietaEarthquake,1989; Alaska Earthquake,1964) haveraised concerns of approaches implemented for thedesign of the deep foundations. As this complex
phenomenon has received considerable attention [12].In the past, the seismic motions have been used as
the input ground motion at the base of thesuperstructure without considering the site effects
and the soil-pile interaction.
The soil-pile-superstructure interaction (SPSI) is
included of the two sources: Inertia and KinematicInteraction. The kinematic interaction is due to thepresence of pile foundation on or in the groundsurface that causes the ground motions deviate fromfree-field notions[3]. Inertia interaction is due to thekinematic
interaction
transmitted
tothesuperstructure.

479

2. SPSI ANALYSIS WITH BNWF
BNWF approach is a simplified method that has
been an old popular and widely used in professional
design offices due to computationally time saving
and accurate. The Winkler’ foundation assumptions
are implemented in the Beam on Nonlinear Winkler
Foundation (BNWF) method. The main assumptionof this method is that the displacement of the eachlayer of soil profile is without dependence on the
responses adjacent layers and consequently theshear stress between soil layers is ignored. The
layered soil deposit can be estimated by thisapproach because the soil properties are allocated todetermine the soil stiffness at the each level. In
addition, the energy dissipation is consideredthrough the material and the radiation damping. The
Soil-Pile behavior can be modeled nonlinearity andlinearity based on solutions envisaged in the time
and the frequency domain. Moreover, thenonlinearity soil behavior, the radiation damping,
the hysteretic damping and gapping can beconsidered in BNWF models.
As Fig.1 shown, in this method, the pile is discrete
to a number of elastic finite beam-column segments
so that the pile mass is modeled as the lumped massin the nodal location. Each nodal is connected to themechanical elements to represent the soil
characteristics. Based on this simulation, the soil
stiffness and the damping are simulated using spring(linear or non-linear) and the damper (radiation andmaterial damping). The springs and dashpots areconnected to nodal locations. There are several
methods for calculation of the spring coefficientssuch as the p-y curves recommended by American
Petroleum Institute (API). El Naggar et al. [8] used
the static p-y curves recommended by O’Neill andMurchison [10] and Matlock [11] in sand andclay(Fig.2).
The variation of the soil parameters affect on thebending moment of the pile (Khari,ISFEHAN). As
regard, the kinematic bending moment at theinterface is affected by the soil nonlinearity behavior
and the frequency content of the seismic motion
even in absence superstructure.
Modeling radiation damping (represent energy loss
due to wave propagation away) is obviously an
important component of the soil-pile-superstructureinteraction analysis so that it is most important in
the far field. Gazetas and Dobry[12] developed aplane-strain model for the radiation damping of thesoil. Based on their research, The radiation-damping
coefficient can be calculated using the following
equations:Fig.1: Schematic view of the proposed modelz>2.5b(2.7)
C r =4Bρ s Vs [1+ (3.4/ π(1-ν )5/4](π/4)0.75a-0.25
z≤2.5bC r =8Bρ s V s (π/4)0.75a-0.25(2.8)

Where ρ s and V s are the density and the shear wave
velocity of the soil, respectively; a is thedimensionless frequency factor (a=2πfB/V s ; B=pile
radius).
Material damping of the soil should be taken into
account when the p-y curves are computed based on
the backbone curve of a secant modulus approach.
The material dashpot coefficient can be determinedby the evaluation of hysteretic damping ratio (β) of
the soil [13].
3. METHODOLOGYANALYSISAND

NUMERICAL

To model of the soil-pile-structure interaction ofabove mentioned, ANSYS code a general-purpose
finite element analysis software was performed. To
evaluate the seismic behavior of the superstructure atransient analysis was implemented under the
dynamic loading shown in Fig.3. A group 3x3 ofsteel piles (diameter= 670mm; thickness= 19mm)
was considered in this study. Each row of this groupwas simulated to a equal pile with the similar480was selected owing to the dynamic parameters of
the soil and the intensity of the seismic excitation.

stiffness. The pile was divided into 13 elastoplastic
elements.

Figure 3- Acceleration time histories and response
spectra at the bedrock roof
The soil profile proposed in the current paper, wasincluded of the two horizontal layers of sand andclay. The upper layer was reconstituted clay(γ=6.68
kN/m3), 6.39 m thick with the plasticity index andthe liquid limit equal to 48% and 80%, respectively.
The lower layer was the sand with 11.1 m thick andthe dry density of 1.66 Mg/m3, D 50 of 15 mm andC u of 1.5. a mass lumped (49100 kg) was attached at
3.8m above the ground surface.

Fig.2: Characteristic shape of p-y curves;(1) sand (2) soft clay (3) stiff clayof “PIPE20” from elements library of above
software. Mass of each part of the pile divided was
lumped at the nodal location using of point elementof “MASS21”. Also, each nodal was connected witha set of the nonlinear spring and the dashpots inparallel on each side of the pile. The soil stiffness
was modeled using the nonlinear spring of“COMBIN39”.Newmark and full Newton-Raphson
method were used to solve the non-linear equationsin the software with amplitude decay factor of 0.005
and to equilibrium convergence, respectively.
The input motion was determined at the pile tip
level using the site response analysis. For thisreason, the nonlinear site response analysis (NERA)

4. ANALYSIS RESULTS COMPUTED
To evaluate the seismic response of the structure,
the acceleration and the displacement time historiesof the structure and the pile kinematic bending
moment within different soil layers can beconsidered. Based on suggestions of some building
codes, the time histories recorded are directly
entered into the fixed base of the structure without
considering the soit effects and the soil-pileinteraction on the time histories recorded. As Fig.4
shown, this suggestion can not be used in the design
and the evaluate of the dynamic response because
the influences of the soil dynamic parameters and481

the kinematic interaction are ignored because thevalue of the peak acceleration is difference with andwithout the site conditions.

Fig.4: Acc. Time histories at
the fixed based structure
the fundamental frequency of the fixed basestructure and the system based on the modal
analysis were equal to 1.94 Hz and 1.32 Hz in thefirst mode as the predominate mode, respectively.
However, the natural frequency of the soil deposit
was calculated based on the one-dimensional wave
equation 0.78 Hz. These frequencies illustrate thatthe soil deposit effects play a very important rule in
the interaction.
The acceleration and the displacement time histories
at the three locations of the system are shown in
Figs.5-6, respectively. A range of material damping
ratios (0%, 5%, 10%) for the pile segments wereconsidered in the numerical dynamic response. As
regards the figures illustrated, with the increase of
the material damping of the pile, the peakaccelerations were decreased.
The frequency content was decreased at the different
levels. So that, the time histories calculated at the
below ground level shows the maximum content.
Also, the maximum acceleration was decreased
along the system height that it can be owing to theenergy dissipation.
The maximum accelerations at the structure level
were 0.0042g and 0.21g with note to the Fig.4 andFig.5(1), respectively. It indicates the soil-pileinteraction has the remarkable effects. In additional,
the Fig. 5(1) shows that the existing of the damping
affect the seismic responses but it is clear that its
effectives is more less when the value of thematerial damping increase.

Fig.5: Acc. Time histories with damping(1) the structure level (2) the pile cap level(3) the below ground level
The influence of the material damping of the pilesegments were more effect on the movement of thesystem. As Fig.6 shown, the displacement time
histories are more different for the damping 5% and10%. Point remarkable in this Fig. can be stated thatthe peak displacement are occurred in the higher
time. The values of the maximum displacements at
the pile cap and the structure level are notnoteworthy.

482

equivalent linear soil. Several aspects will affect theaccuracy of the analysis such as the static p-y curveused to simulate the soil stiffness, not considering of
the soil stiffness degradation, and decay factor in
Newmark method. The results show that the peakacceleration is varied under the kinematicinteraction. In additional, the material damping of
the pile is more effective on the displacement time
histories. However, the intensity of the seismic
motion should be considered in this model.
6. ACKNOWLEDGEMENTS
The research was undertaken with support from
GUP grant (No: 02H09) under the Universiti
Teknologi Malaysia (U.T.M.). The first author
would like to thank the Ministry of Higher
Education (MOHE) and the Research Management
Center (RMC) for the financial supports during thisstudy.
7. REFRENCE
1. Khari, M., K.A. Kassim, and A. Adnan, The
Influence of Effective Confining Pressure
On Site Response Analyses. Asian Journalof Earth Sciences, 2011. 4(3): p. 148-156.
2. Khari, M., A.B.K. Khairul, and A. Adnan,
Evaluation of Kinematic Bending Momentof Piles subjected to Seismic Motions, in
9th International Congress on Civil
Engineering. 2012: Isfahan University ofTechnology (IUT), Isfahan, Iran.
3. Khari, M., K.A. Kassim, and A. Adnan, The
Effects of Soil Model on Site Response
Analyses. Asian Journal of Scientific
Research, 2012a. Under Review.
4. Maheshwari, B.K., et al., Three-dimensional finite
element nonlinear dynamic analysis of pilegroups for lateral transient and seismic
excitations.
Canadian
Geotechnical
Journal, 2004. 41(1): p. 118-133.
5. Wu, G. and W.D.L. Finn, Dynamic nonlinear
analysis of pile foundations using finite
element method in the time domain.
Canadian Geotechnical Journal, 1997.
34(1): p. 44-52.
6. Cairo, R., et al., Soil-pile kinematic interaction:from research to practice. Italian
Geotechnical Journal.
7. Kaynia, A.M. and E. Kausel, Dynamic behaviourof pile groups. Second International
Conference on Numerical Methods in
Offshore Piling, 1982.

Fig.6: Dis. Time histories with damping(1) the structure level (2) the pile cap level(3) the below ground level
5. CONCLUSION
The influence of the soil-pile interaction (so-calledkinematic interaction) on the seismic response of thestructure was investigated in the current paper. A
BNWF method was developed so that the soil
stiffness and the damping (radiation and material)
were considered. A range of the material damping of
the pile segments were carried out. The results ofthis paper are applicable to linear structure and the483

8. model for nonlinear seismic response analysis ofoffshore piles with nonlinear input groundmotion analysis. Canadian Geotechnical
Journal, 2005. 42(2): p. 365-380.
9. Nogami, T., J. Otani, and H. Chen, Nonlinear
Soil-Pile Interaction Model for Dynamic
LateralMotion.
GeotechnicalandGeological Engineering, 1992. 118(1): p.
89-106.
10. O’Neill, M. and J. Murchison, An Evaluation of
P-Y Relationships in Sands. 1983:
University of Houston.
11. Matlock, H., Correlations for Design of
Laterally Loaded Piles in Soft Clay, inProc. 2nd Offshore Technology Conf.
1970: OTC 1024,Houston. p. 577-594.
12. Gazetas, G. and R. Dobry, Horizontal Responseof Piles in Layered Soils. Journal of
geotechnical engineering, 1984. 110(1): p.
20-40.
13. Finn L., W.D., A Study of Piles during
Earthquakes: Issues of Design andAnalysis, in Bulletin of Earthquake
Engineering. 2005, springer. p. 141-234.

484

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Investigation on Finite Element Modeling of Group of Stone
Columns
M. Mirshekari 1, C. Behnia 2 and S. Fakhretaha 3
MSc Student, School of Civil Engineering, University of Tehran, Tehran, Iran2
Ascociated Professor, School of Civil Engineering, University of Tehran, Tehran, Iran3
BSc Student, School of Civil Engineering, University of Tehran, Tehran, Iran1

ABSTRACTIn recent years, stone columns have been widely used to increase load bearing capacity as well as settlement reductionof soft soils. 2D finite element methods are usually used for analysis of load-displacement in the soil reinforced by stone
columns. Plain strain modeling of stone columns is frequently carried out by idealization of equivalent continuous stone
strips. This might result in changing the failure modes of stone columns. For instance, bulging failure mode, one of theimportant failure modes in stone columns, is neglected in the stiff axe of equivalent stone strip. The usual method isused for defining the equivalent stone strip, does not consider the failure mode variations. In this paper a new method is
purposed for 2D modeling of stone columns. Plaxis 2D is used and the analysis is carried out considering various
diameters and spacing of stone columns. Floated and rigid based stone columns have been studied applying the
proposed method. Comparing the results with those obtained by 3D modeling analysis shows the accuracy of the
proposed method.
Keywords: Stone Columns, Finite element, Plain Strain, Equivalent Stone Strip

1. INTRODUCTION
Land scarcity near existing urban areas due to fastdevelopment and urbanization, often necessitates the use ofsome sites in which soil has low quality (such as soft clays).
In this condition, the soil may be unsuitable for short or
long-term construction activities. Hence, the soil properties
must be improved.

Fig. 1. Failure surface assumed by Barksdale andBachus(1983)

Several methods have been presented for soil improvement
such as dewatering, compaction, preloading, grouting, deep
mixing, stone columns, deep densification and soil
reinforcement. Many of these techniques have been used formany years. The use of stone columns as a technique of soil
reinforcement is frequently implemented in soft cohesive soil.
Use of stone columns leads to increase in bearing capacity
accompanied by significant reduction in settlement Using
granular, freely drained material in stone columns,
consolidation settlement is accelerated, and post construction
settlement is minimized.
Theoretical, experimental and field study on behavior ofstone columns have been investigated in last decades. Hughes
and Withers (1974) performed pioneering laboratory studies
of sand columns within a cylindrical chamber containing
clay. They used radiography to track the deformations
occurring within and outside the column. They found that
confining pressure of stone columns represented themeasured column behavior very well. As a result of theirresearchs, they proposed the ultimate vertical stress (q) as afunction of lateral effective stress of stone column. Barksdale
and Bachus (1983) assumed a straight rupture surface for soil
improved by group of stone columns, beneath the foundation.
Their assumption made a triangular failure block, as shown in
Fig. 1. They proposed ultimate bearing capacity of improved
soil, considering straight failure surface.
Priebe (1995) considered the effect of compressibility of thecolumn material and the overburden. He developed design

485

A
S

S

W=AS/S
Fig. 2. Stone Column strip idealization presented byBarksdale and Bachus (1983)
In this paper a new method is purposed for 2D modeling ofstone columns. Plaxis 2D is used and the analysis is carriedout considering various diameters and spacing of stone
columns. The results are compared with those obtained by 3D
modeling analysis.

2. Modeling and analysis

In order to study on the finite element modeling of stone
columns plaxis package is used in this paper. Plaxis package
is one of the most prominent software’s are operate with thefinite element theoretical base. Plaxis package is widely used
for stone columns modeling in the practical projects. Plaxis
2D, version 8.2 is used in this paper and the results arecompared with palxis 3D, version 1.1 in order to define moreaccurate method for 2D modeling of stone columns.

3. Validation of plaxis 3D(PLAXIS 3D) is validated by modeling of a stone column overaxial loading based on load test carried out by Thompson et al.
(2009). The subsurface profile at the project site consists of soft,
nonuniform clay classifying as sandy lean clay (CL) to a depthof
approximately 4 to 5 m. Groundwater was encounteredapproximately 1 m below the ground surface. The clay andgravel properties are assumed exactly as same as which
presented by Thompson et al. (2009). Fig 3. Exhibit acomparison between results of modeling in plaxis 3D and theresults which obtained in full scale load test is carried out by
Thompson et al. (2009).

0

Applied Vertical Stress (kPa)
500
1000

1500

0
Settlement (mm)charts to calculate the settlement of single and strip footing
reinforced by a limit number of stone columns.
Finite element method is recently used to determine thesettlement of a soil improved by group of stone columns.
Plaxis is a common Finite element package used forsettlement determination. Results obtained by finite element
modeling Plaxis have been verified with those obtained byempirical and field studies in several past studies. For
example Ambily and Gandhi (2004) studied a laboratorymodel of single stone column in soil. In a case of static
loading, they validated numerical results obtained by plaxis
with the experimental results obtained by laboratory model.
Barksdale and Bachus (1983) proposed idealized stone strip
method to model group of stone columns as a plain strain
situation. Idealized stone strip is shown in fig. 2.

20
40
60
80

Thompson et al. (2009)This study, Plaxis 3D
Fig. 3. Validation of Plaxis 3D

As could be seen in Fig. 3. the differences between the resultsof plaxis 3D modeling and full scale load test, Thompson(2009), might be neglected. Hence plaxis 3D is validated andcould be used for the verification and précising of plaxis 2Dmodels.

4.

Examples

ρ=

d2
× 100
(1.13S ) 2

The Mohr-Coulomb criteria is used for soil and stone
columns behavior. Also since in the most practical projects,
gravel blanket is used above the stone columns, a gravel
blanket is assumed beneath the load and above the stone
columns. This gravel layer can distribute the stress between
stone columns. In addition, it can dissipate excess pore waterpressure. Thickness of this gravel blanket is 0.5 meter in thismodeling. The columns are placed in square pattern andlength of stone columns varies between 10 to 20 meter indifferent models. Geometry of model in plan and section are
shown in Fig. 4 and Fig. 5. Stone columns diameter are
between 0.8 to 1.2 meter in models. This range is mostcommon diameters of stone columns in practical projects.
The equivalent stone strip thickness is determined withbarksdale and bachus method, as can be seen in Fig. 2.
spacing of stone columns may vary between 2 and 3 meters.
Geometrical parameters of modeling are shown in Table. I.
Each column acts within a cylindrical cell with a radius of
influence denoted by Re (Fig. 6). Balaam and Booker (1981),
related the radius of influence to the actual column spacing by
the relation Re = c.S, whereS is the actual spacing (from centerto center of the columns) and c is a constant having values of
0.525 and 0.564 for triangular and square patterns,
respectively. Considering the radius of influence of stone
columns in square pattern, the area replacement ratio(ρ), total
area of stone columns over area of unreinforced soil, may be
defined as :(1)

By using Eq. (1) the area replacement ratio in different
models varies between 6% to 28%. This range of area
replacement ration may represent different stone columns
geometries.

486

Table I. Geometrical properties of models
Spacing ofStone
Length ofThickness of
Stone
Columns
Stone
GravelColumns (m) Diameter (m)Columns
Blanket (m)
(m)
2
0.8
10
2.5
1
15
0.5
3
1.2
20

Fig. 4. Plan of Stone Columns and The Gravel Blanket

In this modeling, site’s soil is defined purely cohesive andstone column’s material is defined purely frictional material.
The soil is completely saturated and the water level is at the
ground surface. Table. II indicates geotechnical parameters ofsoil and stone column’s material.
Table. II. Site’s soil and Stone Columns specifications, usedin 2D and 3D Modeling
RegionSite’s Soil
Stone Column Material
C (kPa)250040
φ (Degree)
E (MPa)3200
ϒ unsaturated (kN/m3)
16
18
ϒ saturated (kN/m3)
18
20
Distributed loading is assumed and the load acts in a 10*10
meter surface above the gravelly blanket. In reality it canrepresent a liquid storage tank or embankment load.

5. Results

160
140
120
100
80
60
40
20
0

Distributed Load (kN/m2)

Fig. 5. Section of Stone Columns and the Gravel Blanket
Modeled in Plaxis 2D

The results of presented modeling are shown as aload-displacement curve. Since in the practical projects
maximum displacement beneath the foundation is used asabsolute settlement, in this study maximum displacement isused in load-displacement curves. All modeling results aregathered and analyzed. Fig. 7. shows the resultant of thisstudy as a load-displacement curve.

Fig. 6. Definition of radius of influence in square pattern ofstone columns, Balaam and Booker (1981)

2D Results
3D Results

0

20

40

60

80

Maximum displacement (cm)Fig. 7. Load versus maximum settlement in the 2D and 3Dmodels of soil improved with stone columns
As might be seen the maximum displacement of soil in 3Dmodels is higher than maximum displacement in 2D models
487

in light loads, and lower in heavy loads. In fact, however 3D
displacements commence from a higher level, but theelasticity modulus of improved soil is higher in 3D models.
So the displacement in 3D models are lower in heavy loads. It
seems this difference occurs for the variation in failure modeof stone columns. For instance bulging failure mode can notoccurs in the stiff axe of idealized stone strip in 2D modeling.
Also the punching failure mode of stone columns happens
harder in idealized stone strip. Table III shows some results
too.
Table III. Results of 2D and 3D modeling of group of stone
columns
Applied LoadMax Disp. 2D
Max Disp. 3D(kN/m2)
(cm)
(cm)
10
1.46
3.1
50

7.31

9.8

100

21.24

20.4

120

32

29.3

150

56.1

50.5

155

70

66.1

Considering the results, it seems 2D modeling results can be
improved using Eq. 2.

S 3 D = S 2 D ×1.3 :qqultimate

S 3 D = S 2 D × 0.9 :qqultimateacknowledgment is given to Prof. Ali Fakher for his
astutely advices.
8. REFERENCES[1] Ambily, A.P. and S.R. Grandhi, 2007. “Behavior ofstone columns Based on Experimental and FEM
Analysis”, Jour. Of Geotechnical and Geoenviromental
Engineering. ASCE, 133(4): 405-415.
[2] Barksdale, R.D. Bachus, R. C., “Design and constructionof stone column”, Vol.1[3] Hughes, J.M.O. and Withers, N.J. (1974). “Reinforcingof Soft Cohesive Soils with Stone Columns”, GroundEngineering, Vol.7, No.3, 42-49.
[4] Hughes, J.M. O., N.J. Withers, and D.A. Greenwood,
1976. “A field trial of reinforcing effect of stone column
in soil”. Proc., Ground Treatment by Deep Compaction,
Institution of Civil Engineers, London, 32–44.
[5] Priebe, H.J., 1995. “The design of Vibro replacement.
Ground engineering”, 31-37.
[6] Thompson, M. J., Wissmann, K. J., Pham, H. T. V.,(2009), “Performance Monitoring of a Rammed Aggregate
Pier Foundation Supporting a Mechanically Stabilized EarthWall”, Journal of performance of constructed facilities,
July/August,2009, P 244-250.

≤ 0.5
(2)

> 0.5

Which
S 2D and S 3D , are the settlement in 2 and 3 dimensionsq, is the available stress leveland q ultimate , is the bearing capacity of improved soil.
The Eq. (2) may be used to precise 2D modeling results,
where plain strain model is used.

6. Conclusion
In this paper, a more accurate method is proposed todetermine maximum settlement of soil which is improved
with stone columns. Stone columns with different spacing’s,
diameters and lengths are modeled and the load-displacement
curve is illustrated for the resultant of models. Differences
between 2D and 3D results are seen and it seems that thesedifferences have been occurred because of the changing in
failure modes of stone columns. Finally by comparing 2D and
3D models results 2 coefficients are proposed to precise 2D
finite element modeling results.
7.

ACKNOWLEDGMENT
Acknowledgment is given to my dear friend Miss.
Sanaz Mahmoudpour, for her kindly supports. Also,
488

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Suction Controlled Triaxial Apparatus for Saturated-Unsaturated Soil Test
Luky Handoko, Noriyuki Yasufuku, Kiyoshi Oomine, Hazarika Hemanta
Department of Civil Engineering, Kyushu University, Japan
ABSTRACT
A suction controlled triaxial apparatus has been developed in Geotechnical Engineering Laboratory of Kyushu
University. The apparatus has been designed to obtain mechanical properties of unsaturated soil such as shear strengthand stiffness, as well as hydraulic properties of unsaturated soil, i.e. soil water retention curve by measuring drainedwater volume change during applied matric suction. Not only for unsaturated condition, this apparatus can be also used
to conduct conventional test on saturated soil. This paper focuses on the performance of the apparatus, by performing aseries of multistage test on unsaturated soil under drying and wetting path. The hydraulics and mechanical properties of
unsaturated soil are presented and discussed.
Keywords: triaxial apparatus, unsaturated soils, suction control, hydraulics and mechanical properties

1. INTRODUCTION

2. SUCTION CONTROLLED TRIAXIAL APPARATUS

Stress state variable, such as net normal stress and matricsuction has been successfully governing behavior of
unsaturated soils [1]. Those variables have significant
influence on hydraulics and mechanics properties. Many
researchers have been focusing study on determining shear
strength [2]-[4], small strain stiffness [5]-[7], and soil water
retention curve [8] –[9] of unsaturated soil through laboratory
experimental. To obtain hydraulic and mechanical propertiesof unsaturated soil, an accurate control and measurement
equipment is needed. Triaxial apparatus for unsaturated soil
has been developed by many researchers [10]-[13].
Due to climate change, unsaturated soil prone to havehysteresis features and should no longer be ignored [14]. To
investigate the hysteresis effect on unsaturated soil hydraulics
and mechanics properties, an apparatus which able to control
suction needs to be developed. This paper presents thedevelopment of suction controlled triaxial apparatus which
can be used to conduct test on unsaturated soils as well as
conventional saturated condition. The apparatus has
independent control of cell pressure, pore air pressure andpore water pressure, equipped with Local Deformation
Transducers (LDTs) to measure strain at small range andPressure Differential Transducer to measure both drainedwater volume change and total volume change. With thoseequipment, it is possible to obtain both hydraulics andmechanics properties of unsaturated soil in a series ofmultistage test. Five centimeter diameter and ten centimeterheight of silty sand specimen was used to conducted
multistage triaxial test on different matric suction and stresspath. Hydraulic and mechanical properties of the specimen
are going to be presented and discussed.

A suction controlled triaxial apparatus has been developed in
Geotechnical Engineering Laboratory of Kyushu University
collaborating with Geo-Research Institute, Kobe, shown in
Fig. 1. This apparatus can be used to perform test underunsaturated condition by using axis translation techniquecontrolling both pore water and air pressure to maintainmatric suction within the specimen, as well as under saturated
condition by providing a channel to apply back pressure.
Many instruments for control /measurement purposes areequipped in the apparatus, such as: external and internal loadcell to measure deviator load; pressure transducer to measurecell pressure, pore water pressure at bottom of the specimen
and pore air pressure at the top of specimen; Pressure
Differential Transducer to measure drained water volume
changes and total volume changes, LVDT to measure axial
large strain range externally, LDTs and Pi-gauges to measure
axial and radial small strain range internally.
Pressure control/measurement
Air pressure is supplied at a constant value about 700kPa to
the pressure regulator. From pressure regulator, it is divided
into four channels for controlling cell pressure, air pressure
and two channel of pore water pressure independently.
Pore air pressure is applied at the top of the specimen
through normal porous disk, while pore water pressure isapplied at bottom through high air entry ceramic disk. The air
entry value of ceramic disk is about 3 bars (300 kPa). High air
entry ceramic disk was glued and completely sealed at thepedestal and connected with water channel at the bottom. At
the middle of ceramic disk, metal porous disk is installed,
mainly used during saturating the specimen or during
saturated triaxial test. Fig. 2 shows the schematic diagram ofpressure regulation in triaxial apparatus.

489

a

b

Volume change measurement
Total volume change and drained water volume change canbe measured by using pressure differential technique [15].
Open ended inner cell is used to measure total volume changeof soil. The fluctuation of water level inside the inner cell
represents the change of total volume change. Drained water
volume change is calculated by measuring the amount ofwater drained in or out of specimen. For complete descriptionof pressure differential technique for measuring volume
change please refer [15].
Strain Measurement
During shearing, strain of deformed specimen is measured
externally and internally. Linear Variable Differential
Transformer (LVDT) (range 0-20% strain) is placed outside
the chamber to measure large strain range of specimen during
shearing. Internally, LDTs [16] are attached at the specimen
to measure axial strain, while other instruments so called
pi-gauge are attached horizontally to measure lateral strain.
The maximum displacement can be handled by LDT is about
1.5 mm (1.5% of strain). These LDTs and Pi-gauges were
produced by Geo-Research Institute, Kobe.

c

3. TEST METHOD

d

3.1 Materials

Fig. 1 Suction controlled triaxial apparatus, a) schematic
diagram, b) photo, c) normal porous disk (top of specimen)
and d) high air entry value disk and porous metal filter(bottom of specimen)

Two types of soil were used in the tests; they are low
plasticity silt called Red Soil from Okinawa Perfecture, Japanand sand soil called Toyoura sand. The grain size
distributions are shown in Fig. 3 and the basic properties areshown in Table 1. Red soil was prepared to conduct soil water
retention test, while Toyoura sand was prepared for shearing
test. Specimen dimension prepared in the apparatus is 10 cm
in height and 5 cm in diameter for shearing test. To reduce thetime consumed in soil water retention test, specimen height
was reduced to be 5 cm height, while the diameter is same.

Fig. 2 Schematic diagram of pressure regulation
490

The specimen was prepared by using moist tamping
method compacted at a split mold attached to the pedestal of
triaxial apparatus [17]. Required amount of dry soil was
mixed with predetermined amount of de-aired water and thenkept about 16 hours before used. Internal split mold was used
to make the specimen, attached at the base pedestal of theapparatus. The specimen was compacted by 5 equal layers forsoil water retention test and 10 equal layers for shearing test.

Net normal stress variable was set to constant value at 20kPa.
Table 2. Applied stress and stress state variable
Stages u a (kPa) u w (kPa)u c (kPa)u a -u w (kP)
I
25
25
45
0
II
30
25
50
5
III
40
25
60
15
IV
30
25
50
5
2 RESULT AND DISCUSSION
Among of equipment mentioned above, LDTs performance
may change with their use, so calibration is needed after used
few times. For use in multistage test, LDTs will experienceloading and unloading of deviatoric stress for many times in
one series of test. This equipment needs to be calibrated
properly before used by using digital micrometer as shown in
Fig. 4.

Fig. 3 Grain size distribution of Red soil and Toyoura sandTable 1. Basic properties of Toyoura sandSoil type G s
e
w (%)
γd
Red soilToyoura

2.64
2.65

0.63
0.68

10.6
10

3

(kN/m )

16.23
15.8

γ b(kN/m 3 )
17.95
17.38

3.2 Test Procedures
Reference [18] explained the multistage testing procedure intriaxial apparatus by increasing soil suction of unsaturated
soil. This procedure is mainly used to provide unsaturated soil
properties of different applied matric suction and different
path by using single soil specimen. Even though it is possible
to obtain hydraulics and mechanics properties by using single
soil specimen, in this test, two different tests were conducted
to evaluate the performance of the apparatus. Water retention
test and shearing test were conducted using the apparatus.
For water retention test, the specimen test was started fromsaturated condition. Pore air pressure and cell pressure wasincreased in the same value to maintain destined matricsuction and to keep constant of net normal stress. Pore waterpressure was set constant about 25kPa. Net normal stress was
also set constant about 50kPa. Matric suction was varied to
3kPa, 10 kPa, 30kPa, 100kPa, 170kPa, 250kPa, 100kPa and
30kPa.
For shearing test, consolidated drained condition waschosen with shear rate about 0.01 mm per minutes. The
specimen was tested initially from saturated condition. Afterbeing saturated, the specimen was sheared until reaching peakvalue and then unloading. After shear loading and unloading
was finished, consolidation for next stage predeterminedmatric suction started, and then sheared again after reachingequilibrium state. This condition was repeated for next stage.
A typical set of stresses for each stages are given in Table 2.

Fig. 4 LDT calibration techniques
Calibration for LDT is polynomial order two, while pigauges follows linear curve. Fig. 5 shows the repeatedcalibration by loading and unloading for 2 times. It shows thatLDT and pi gauges are still in good calibration due torepeated loading. The calibration curve does not change somuch, and expected to work properly during multistage test.

491

a)

b)

Fig. 5 Repeated calibration for a) LDT and b) pi-gauges
Soil Water Retention Test
Fig. 6 shows the drained water evolution versus time. Itshows that drained water volume change can be monitored
well. The result shows that red soil reaches equilibrium timeabout 8-10 days. After finish all consolidation step, watercontent of specimen was measured, and then recalculatingwater content for each applied matric suction by backcalculation. As the result, SWRC of red soil can be drawn in
Fig. 7. This figure shows the hydraulic hysteresis of red soil.
At the same matric suction, wetting process will give lower
water content than drying process. Environmental conditionof specimen during soil water retention test is more similar
with field condition by applying radial pressure (cellpressure) at the specimen. This is one of the advantages ofusing triaxial apparatus to measure soil water retention curve.

using hand, and may unstable. Reference [16] suggests a
fixing device to attach the LDT, so it can be aligned properly.
Expected correct line is drawn in red line. Smaller strain
range (below 0.01%) of stress-strain diagram and related
elastic moduli at small strain are presented in Fig. 9 and Fig.
10. These figure more clearly show the stiffness of soil at
small strain range. It shows that in small strain range, thestress-strain relationship can be represented by linear line.
Elastic moduli of soil changes depend on the matric suction
applied.
Stage III

Stage IV

Stage II
Stage I

Expectedcorrectline

Fig. 8 Stress strain relationship of 4-stages multistage triaxial
test

Fig. 6 Monitored drained water volume change

Fig. 9 Stress-strain diagram in small strain range of each stage

Fig. 7 SWRC experiment data fitted with Van Genuchten
fitting curve
Shearing test
The stress-strain relationships for 4-stages (different matricsuction) shearing test is shown in Fig. 8. It shows that thestrength of soil increases as the increasing matric suction.
And also, the curve is steeper as the increasing matric suction,means that stiffness is also increasing. At the early shearing
test (matric suction = 0kPa), the graph seems to be incorrect.
This problem may occur when the LDT was not attachedproperly. In this test, the LDT were attached manually by492

Fig. 10 Elastic moduli of each stage
3 CONCLUSION
A suction controlled triaxial apparatus has been developed in
Geotechincal Engineering Laboratory of Kyushu University
collaborating with Geo-Research Institute, Kobe. The
apparatus has an automatic system to collect and store thedata. It can be used to conduct multistage test to obtainhydraulic (SWRC) and mechanics (shear strength andstiffness) properties of soil. Two different series of multistage
test has been conducted to evaluate the performance of theapparatus. From the experimental test result, it can beconclude that: the developed suction control apparatus is able to provide
good monitoring and storing data automatically,
 the apparatus equipped with LDTs is to measure small
strain range and works well based on the test result, unsaturated soil is having hysteresis features due todifferent process (drying and wetting). This apparatus isable to observe this phenomenon. It can be shown
clearly in the hydraulic properties (SWRC) test result.
4 ACKNOWLEDGESThis research was financially supported by Grant-in Aid forscientific research (A) No. 22246064 from JSPS lead by N.
Yasufuku. Authors also gratefully acknowledge support from
Geo-Reserch Institute for the good collaboration to develop
the apparatus.
5 REFERENCES[1] Fredlund, D.G., and Morgenstern, N.R. Constitutive
relations for volume change in unsaturated soils,
Canadian Geotechnical Journal, 1976, 13(3): 261–276[2] Escario, V. and J. Saez . "The Shear Strength of Partly
Saturated Soils." Geotechnique, 1986, 36(3): 453-456.
[3] Rassam, D. W. and F. Cook. "Predicting the Shear
Strength Envelope of Unsaturated Soils." Geotechnical
Testing Journal, 2002, 25(2): 215-220.
[4] Guan, G. S., H. Rahardjo, et al. "Shear Strength
Equations for Unsaturated Soil under Drying andWetting." Journal of Geotechnical and Geoevironmental
Engineering, 2010, 136(4): 594-606.
[5] Mendoza, C. E., J. E. Colmenares, et al. Stiffness of an
Unsaturated Soil Mechanics. Advanced Experimental
Unsaturated Soil Mechanics. A. Tarantino, E. Romero

and Y. J. Cui. Trento, A. A. Balkema Publishers, 2005,
199-204.
[6] Mendoza, C. E. and J. E. Colmenares. Influence of theSuction on the Stiffness at Very Small Strains.
Proceedings of the Fourth International Conference on
Unsaturated Soils. G. A. Miller, C. E. Zapata, S. L.
Houston and D. G. Fredlund. Arizona, ASCE, 2006.
[7] Nyunt, T. T., E. C. Leong, et al. "Strength andSmall-Strain Stiffness Characteristic of Unsaturated
Sand." Geotechnical Testing Journal, 2011, 34(5).
[8] Tami, D., H. Rahardjo, et al. "Characteristics of Scanning
Curves of Two Soils." Soils and Foundations, 2007,
47(1): 97-108.
[9] Liu,Q., et al. Automatic soil water retention test systemwith volume change measurement for sandy and silty
soils. Soils and Foundation, 2012, 52(1): 368-380.
[10] Aversa, S. and M. V. Nicotera. "A Triaxial andOedometer Apparatus for Testing Unsaturated Soils."
Geotechnical Testing Journal, 2002, 25(1): 3-15.
[11] Lauer, C. and J. Engel. A triaxial device for unsaturated
sand — New Developments Unsaturated Soils:
Experimental Studies. T. Schanz, Springer Berlin
Heidelberg, 2005, 93: 301-314.
[12] Cabarkapa, Z. and T. Cuccovillo. "Automated Triaxial
Apparatus for Testing Unsaturated Soils." Geotechnical
Testing Journal, 2006, 29(1): 21-29.
[13] Padilla, J. M., W. N. Houston, et al. An AutomatedTriaxial Device for Unsaturated Soils. Proceedings of the
4th International Conference on Unsaturated Soils.
Arizona, ASCE, 2006, 1775-1786.
[14] Li, X. S. Modelling of Hysteresis Response for Arbitrary
wetting/drying Paths. Comput. Geotech, 2005, 32, No. 2,
133–137.
[15] Ng, C. W. W., L. T. Zhan, et al. "A New Simple System
for Measuring Volume Changes in Unsaturated Soils."
Canadian Geotechnical Journal, 2002, 39: 757-764.
[16] Goto, S., F. Tatsuoka, et al. "A Simple Gauge for Local
Small Strain Measurements in the Laboratory." Soils andFoundations, 1991, 32(1): 169-180.
[17] Ladd, R. S. "Preparing Test Specimens Using
Undercompaction." Geotechnical Testing Journal, 1978,
1(1): 16-23.
[18] Ho, D. Y. F. and D. G. Fredlund. "A Multistage Triaxial
Test for Unsaturated Soils." Geotechnical Testing
Journal, 1982, 5: 18-25.

493

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Monitoring Program and Dilapidation Survey for
Grouting of Karst Cavities
Hasan A. Kamal and Hassan J. Karam
Kuwait Institute for Scientific Research, Kuwait

ABSTRACT
A total of eight ground-surface subsidence were detected in a residential suburb. The subsurface failure cases in thesuburb were investigated in detail after their development. Results of the investigations indicated that the geological
profile in this suburb consists of overburden soil which is underlain by limestone bedrock. It was concluded thatraveling of the overburden soil into the underlying Karst cavities was the main cause of ground-surface subsidence
development. Filling the Karst cavities with cementatious grout was chosen as a treatment measure for the cavity
problems. Before and during the grout works for filling the Karst cavities, continuous monitoring of elevations in someselected points at critical locations were considered an essential tool to detect any slight movement or change that may
be evaluated and considered a warning for larger movements or cavity and ground-surface subsidence formations. The
continuous monitoring program was based on two major elements; dilapidation survey of selected high risk locations
and the measurement of water levels for determining water fluctuations. Dilapidation was achieved throughtopographical site surveys including houses, utilities and general surveys along with a set of pictures illustrating
important features. Station points were set all over the site and their elevations were checked regularly. Regularperiodic evaluation of elevations, water-table levels and crack-meter readings were used as tools to evaluate the impactof the grouting works. Status of existing structures was checked regularly using crack-meters along with water-tablelevels through available piezometers. This paper contains description of the conducted dilapidation and topographicalsurvey. Results of the monitoring program of the grouting works and their impact on the ground-surface subsidence isalso presented in this paper.
Keywords: subsidence; grout; Karst; soil; topographical

1. INTRODUCTION
A number of ground surface subsidence in the form ofsinkholes of various sizes had been occurring in a
residential suburb in State of Kuwait [1]. The first sinkholeof April 1989 (SH-1) was immediately backfilled withgravel, sand, and concrete. The occurrence of the second
and third sinkholes during the same months led to thedecision to evacuate more than 130 surrounding houses [2].
Extensive investigation studies had been conducted in thisaffected residential suburb in order to understand the causesof those sinkholes and recommend appropriate treatment
methods [2]-[4]. The studies that included topographic andgeophysical surveys, geological, and geotechnicalinvestigations revealed the existence of underground Karst
cavities within the Limestone bedrock underlying theoverburden soil caused the sinkholes.
The studies also reported that the geological profile of thisresidential suburb consists of 35 to 40 m thick overburden
soil comprising dense to very dense predominantly quartz
sand, overlying the Karst limestone bedrock. The cause of
the sinkholes had been attributed to the dissolution of thelimestone bedrock and subsequent raveling of theoverburden soil cover into the underlying Karst cavities.
Accordingly, treatment methods were recommended tominimize the recurrence of the sinkholes composing offilling the underground cavities by injecting cement grout
at depths varying between 32 to 60 m in a pilot area and byeliminating the possibilities of collapse of the upper level
Karst cavities within the limestone bedrock [5].
The pilot area located in Sector A1 of the residential areaclassified as the evacuated area (Fig. 1), was divided into
six treatment zones, TA1 to TA6, for sequential injection ofgrout according to these zones. In general terms, thegrouting program work required performing dilapidation
survey, drilling exploratory holes with in situ testing andsampling; designing a grout mix, and developing quality
assessment program; filling in of deep limestone cavities
using cement-based mortar; permeation grouting of
remaining deep voids and fractured rocks using
cement-based grout; drilling control holes at the end ofgrouting and overburden treatment.
In this paper, the monitoring program used during thetreatment implementation methods is described. The
continuous monitoring program focused on the impact of
the treatment drilling and grouting works on the existing
structures on the site, ground surface movement andfluctuation of water-table levels. The results are presented
and discussed and conclusions are illustrated.
2. METHODOLOGY OF MONITORING PROGRAM
The purpose of the monitoring program is to monitor the
potential impacts of drilling and grouting works for the
implemented treatment methods on the existing structuresand ground surface condition.
The monitoring program focused on the sinkholes and their
behavior during and after the period of treatment494

implementation works to detect impacts resulted from thetreatment campaign. The following subsections describe
the methods used for monitoring the treatment effects in thearea under study.

Fig. 1 The six treatment zones, TA1-TA6, in the pilot area.
2.1 Dilapidation Survey
Dilapidation survey is a detailed record of land andbuildings at a specific point in time. The survey started
before implementing the treatment methods by describing
the condition of the existing structures, indentifying any
defects, and a schedule of pictures for existed defects.
Large numbers of pictures were taken including generalviews of the area, description and close up photos of someparts of the structures and utilities (Fig. 2). Detailed
identification of location, date of picture taken, image
number, and location on key map were also recorded foreach photo.

The topographical survey concentrated mostly on regularrecording of elevations inside and around the treated areas.
Elevation points were distributed on the ground surface
forming a grid of less than 50 m spacing, in addition topoints selected in each housing unit. The readings were
evaluated and recorded using a total-station surveying
instrument. Any signs of undesired ground movement weredetected and projected into maps and tables.
2.3 Crack Propagation
Crack-meters of an accuracy of 0.001 mm were used torecord movements of selected elements (Fig. 3). Crack
meters were located on site mostly on structures, aroundsinkholes and other sensitive areas suspected withoccurring movement. A number of crack-meters were set
on columns, slabs, and on structural joints on ground andupper floors. One crack meter was set on the asphalt close
to the progressing sinkhole SH-1. Each crack-meter was
identified by a label with all necessary information on thelocation and the structure on which it is set.

Fig. 3. A crack-meter sample used to monitor movements.
2.4 Water-Table Fluctuation
Evaluation and monitoring of groundwater and water-tablelevels is essential for detecting changes or risks, as thewater content affects the soil behavior and stability. Threepiezometers were drilled and were used to evaluate
water-table levels since 2002.
These holes were
unfortunately blocked during the heavy work on site.
Three new holes were drilled and are currently being used
to evaluate water-table levels and detect groundwaterfluctuations. The locations of those new piezometers areshown in Fig. 4.

Fig. 2. Dilapidation picture in area TA1.
2.2 Topographical Site Survey
At the beginning of the monitoring program, a site surveyof the area under study was carried out. The output resultsof this part were drawings for locations of all existingbuildings and facilities such as structures, fences, walls,
roads, sidewalks, over and under-passes, gardens, open
yards, trees, ditches, cannels, culverts and wells, services
and related structures (both above and below ground) suchas water mains, sewers and electricity and telephone poles.
They were considered as the “As-Built” drawings.
495

Fig. 4. Location of the piezometers.

2.5 Visual Inspection
A visual inspection was performed to document thecommon defects, whether they are due to demolishing,
aging, construction etc. A sample picture of those defects isshown in Fig. 5. Those cases pictures reflect the general
status and condition of the site and structures.

therefore; points are concentrated on the houses and aroundthe old sinkhole locations. Fig. 6 shows the selected
monitoring points and their elevations, and Fig. 7 shows
sample records for selected points in zone TA3 whichindicates no significant changes in the elevation of thosepoints during the treatment implementation period.

Fig. 6. Coordinates and elevations of selected points.
68.700Fig. 5. Exposed corroded reinforcement and expelledconcrete at the bottom of a ground floor column.

67.700

3 RESULTS AND DISCUSSION

66.700In the monitoring program, effects of drilling and grouting
on ground surface and existing structures were considered.
The drilling is carried out with a drilling fluid, either waterwith polymer or air alone. The fluid has an ability topermeate more easily in the underground channel because it
is light and not loaded with too much particles. Two typesof grouting treatment were used which included mortar
filling and permeation grouting.
Cavity filling with mortar was carried out as specified by
the gravity process and, consequently, the pressure exerted
during the process was not very high (~ 5 bars). Therefore,
there was no change was recorded in any elevationconsecutive to the cavity filling.
A higher pressure of 30 bars was applied during thepermeation grouting stage. No perceivable changes in theelevation of the surveyed points were recorded as a result of
the permeation grouting. This indicates that there wassufficient room for the voids to redistribute in the groundwithout impacting the surface.
Dilapidation pictures show general views of the area,
description and close up photos of some parts of the villas,utilities, etc. Detailed identification of location, date ofpicture taken, image number, and location on key map arepresented as well on each photo.
For the topographical survey , appraisal of the readings
recorded for elevation points by a total station wasmonitored through any sign of ground undesired
movement. Safety of the housing units is the mostimportant issue before and after treatment implementation,

65.70064.70063.7004-Sep-06

18-Sep-06

2-Oct-06

16-Oct-06

30-Oct-06

13-Nov-06 27-Nov-06

11-Dec-06

25-Dec-06

8-Jan-07

22-Jan-07

5-Feb-07

19-Feb-07

5-Mar-07

19-Mar-07

Fig. 7. Display of elevations for area TA3.
For crack propagation monitoring, a number ofcrack-meters were mounted at different locations andreadings were taken periodically. Fig. 8 is a plan view of
the treated area where the rectangular red marks andnumbers show the locations of the mounted crack meters.
Table 1 is a summary of the accumulated readings of asample of crack-meters.
The amplitude recorded varies from 0 to 0.5 mm for the
typical cracks and 0-3 mm for the dilatation joints. It may
be said that these recorded data are expected and could be
attributed exclusively to the thermal effect. The readings
showing the more significant displacements are recorded
along the dilatation joints at upper floor during thewintertime when thermal effect is measured. The crack
meters have confirmed the fact that there was no impact
recorded from the general survey.
For the water-table level monitoring, the ground water andits effect on the soil is of real importance in evaluating thetreatment method. The water-table level was targeted since496few years before the treatment implementation. Table 2
lists sample reading for the water-table level for the new
piezometers. In general, slight rise in the water-table level
was observed during the treatment implementation
campaign.

Fig. 8. Locations of crack meters.
Table 1. Summary of Crack-Meter Readings
No.

Min.

Max.

(mm) (mm)Difference(mm)

Location

Zone
(TA)

Unit

Floor

Unit

4 CONCLUSION
The outputs of the pilot area under study mainly depends onthe geological profile encountered and the type of the works,
the porosity of the sand. The thickness of the overburden isof 30 meters which here is sufficient to allow redistributionof the porosity during the grouting and the rock grouting
took place with a rather limited pressure.
The monitoring and surveying campaign represented an
important essential component of the treatment
implementation process of underground cavities in the area
under study. The area was intensively treated using two
different types of rock treatments; cavity filling andpermeation grouting.
Regular periodic evaluation of elevations, water-tablelevels and crack-meter readings were used as tools toevaluate the impact of the drilling and grouting works of thetreatment implementation program. During the treatment
implementation using low and high pressure grouting, noalarming signs or significant move or settlement wasobserved due to the grouting works. It signify thatsufficient space for the voids were available to redistributein underground level without impacting the surface.
Continuous monitoring was recommended to be achievedby regular evaluation of elevations in selected higher risk
locations reporting any alarming changes, along with
monitoring water-table levels through available
piezometers in order to assure the safety of the area and its
occupants.

Adjacent

1

0.50

2.70

2.20

3

9

1st

11

2

-3.00

1.00

4.00

3

9

1st

11

3

1.50

1.50

0.00

1

19

GF

-

4

0.00

1.50

1.50

1

19

1st

17

5

2.00

2.00

0.00

1

23

1st

25

6

0.50

1.00

0.50

1

23

1st

-

7

0.00

0.00

0.00

3

Asphalt

-

-

8

0.00

0.00

0.00

3

A2

1st

-

9

0.50

0.50

0.00

3

8

1st

-

10

0.00

0.00

0.00

3

11B

1st

13

11

0.00

0.00

0.00

2

35

1st

36

12

0.00

0.00

0.00

4

63

GF

65

13

-0.50

-0.50

0.00

4

3B

1st

-

14

0.00

0.00

0.00

4

5

1st

7

15

-0.50

-0.50

0.00

1

40

1st

42

Table 2: Sample records for the water-table level

5 ACKNOWLEDGMENT
The authors would like to express their appreciation to the
Public Authority for Housing Welfare (PAHW) and Kuwait
Institute for Scientific Research (KISR) for their facilitativeand support of this project.
6 REFERENCES[1] Abdullah W., and Kamal H., “Characterization ofDesert Karst Terrain in Kuwait and the Eastern
Coastline of the Arabian Peninsula,” in Proc. Sinkholes
and the Engineering and Environmental Impacts of
Karst, San Antonio, Texas, 2005, pp. 35-45.
[2] Al-Rifaiy A., “Land subsidence in the Al-Dhahar
residential area in Kuwait: a case history study," Q. J.
of Eng. Geol. London, vol. 23, 1990, pp. 337-346.
[3] Abdullah W., Mollah M., Al-Mutairi N., Al-Fahad F.,
and Mussallam H., “Evaluation and Treatment ofUnderground Cavities at Al-Dhahar Area”, Vol. II,
Subsurface Exploration Study in the Al-Dhahar Area,
1998, Final Report, Kuwait Institute for Scientific
Research.
[4] Al-Mutairi N., Eid W., Abdullah W., Misak R., Mollah
M., Awny R, Al-Fahad F., “Evaluation and Treatmentof Underground Cavities at Al-Dhahar Area”, Vol. I,
Evaluation and Treatment of Subsurface Conditions at
Al-Dhahar Area, 1998, Final Report, Kuwait Institutefor Scientific Research.
[5] Kamal H., and Abdullah W. “Sinkholes’ Detection in a
Residential Area in the State of Kuwait”, Arabian
Journal for Science and Engineering, accepted, 2011.
497

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Numerical analysis of soil nail walls under seismic condition for different
geometry of the nails
Siavash Zamiran1, Hamidreza Saba2
Department of Civil Engineering, Science and Research Branch, Islamic Azad University, Iran,zamirans@gmail.com
ABSTRACT
2
Amirkabir University of Technology, Tehran, Iran
P

1
P

P

P

P

P

P

In this paper, numerical simulations of soil nail walls under simulated vibrational input have been carried out and theresults are compared with the function of soil nail walls under ordinary statistical loading. The behaviour of geometry ofnails is mentioned under static and seismic analysis. The analysis is carried out with finite difference software called
FLAC. The results are prepared as lateral displacement of the walls and normalized maximum tensile forces for nails.
These results can demonstrate the behaviour of external and internal resistance of soil nail walls under dynamic andstatic analysis. The deformation of wall under static and dynamic manner varies in a wide range. On the other handtensile loads that are produced in nails under static manner are namely 50% less than the dynamic manner.
Keywords: Soil Nail Walls, Dynamic Analysis, Finite Deference Analysis, FLAC

1. Introduction
The behaviour of soil nail walls is the function of soil and nailstrength parameters, geometry of wall, etc. Numerous
investigations are carried out to demonstrate the static andseismic behaviour of soil nail walls. However, with respect todifferent parameters that influence on the behaviour of thesesystems, more investigations about soil nail walls arenecessary. In this research, deferent geometry parameters ofsoil nail walls are mentioned in static and dynamic behaviourof soil nail systems. The strength function of the wall underManjil earthquake acceleration (Iran 1990) loading condition
is mentioned by wall deformation and vibration of soil nailforces.
Mohr-Coulomb constitutive model is selected for numerical
study in this research. The strength function of nail andshotcrete is considered as elastic. The finite deference
software called FLAC is used for numerical study. FLAC canmodel the interaction between nail and soil properly as staticand dynamic analysis in the soil medium.

Thompson and Miller measured the highest nail forces by
inclinometer and calculated the values by the finite-element
method. A comparison of maximum nail forces in measured
values and finite-element calculations in each row is shown in
Fig. 2 with the values that gain from finite deference method,
FLAC. The results show reliable convergence among
calculated values, FEM and FDM. Fig. 3 shows the reliable
convergence of the places that maximum tensile loads occur
through the nail lengths.

2. Verification of the finite deference model
Thompson and Miller (1990) investigated the Seattle’s firstnailed walls. For this vertical soil nail wall, nails were mostly
installed at 1.8m spacing horizontally and vertically. The naillength is 10.7m, except for the length of the top row which is
9.8m. The diameter of the drilled holes is 203mm. Nail bars
are installed at an inclination of 15 degrees, though the firstrow on the high wall was installed at 20 degrees to avoid
utilities. A typical section of the high wall is shown in Fig. 1,[1]. In this research, the stability of this soil nailed vertical cut
is analysed by FLAC.

Fig. 1) Typical section of Seattle’s wallFig. 2) Comparison of maximum nail forces498

Fig. 4) Idealization of the soil nail system.
Fig. 3) The places that maximum tensile loads occur throughthe nail lengths.
3. Numerical modeling considerations
In this study, a 9 meters soil nail wall is considered. 9 rows ofnails with the length of 9 meters are selected for stabilizing
the wall. The selection of the first geometry is based on the
FHWA design booklet. Nail inclination in the first model is
10 degrees, and the horizontal and vertical space between
nails is 1 meter.
Static equilibrium of the wall is obtained with step by step
excavations. Soil mass is modelled in the software by popularconstitutive model, Mohr-Coulomb. Elastic modulus of soil is
20 Mpa; Poisson's ratio is 0.3; unit weight is 17.6 kN/m3;
internal friction angle is 31 degree and cohesion strength is 20
Mpa.
Cable structural element is used to modeling the nail. Also
beam structural element is used to simulate shotcrete innumerical models.
According to some researchers like Wei and Cheng, a thinlayer of material with a thickness of 4.0 mm surrounding thenail is used to model the shearing zone between the nail andthe soil that is shown in Fig. 4. [2]The shear stiffness of the shear zone (grout), Kg, can be
estimated in Eq. (1).

Kg =

2π G
10 ln(1 + 2tD)

(1)
Where G is shear modulus of the shear zone and identical toshear modulus of soil. D is drilled hole diameter and t isannulus thickness of the shear zone and is considered equal to0.004 m. The shear zone cohesive strength (slider) per unit
length can be estimated as Eq. (2).

=cg π c( D + 2t )(2)

Where c is cohesion of soil. Shear zone friction angle isnearly equal to friction angle of surrounding soil.
The properties of cable and beam elements for nail andshotcrete is shown in Table 1.
Earthquake loading is based on shear stress history that is
applied to the base of the model. Quiet boundary should
adjust for bottom boundary of the model. To avoid rejectionof the earthquake wave to the model, free field boundary isused for left and right side of the model.

Table 1) Cable parameters for simulation of nailparameter
value unit
Nail diametermm
Drill hole diameter
100
mm
Young’s modulus of nail200
GPa
Young’s modulus of grout
22
GPa
Young’s modulus of grouted nail29.12 GPa
Annulus thickness
0.004 mNail shear zone cohesive strength
6.28
kPa
Nail shear zone friction angle
31degreeNail shear stiffness of the shear zone
219.6 MPa
Compressive yield strength of the grouted 30
MPanail
Tensile yield strength of the grouted nail30
MPadensity of grouted nail
2200
kg/m3
Young’s modulus of shotcrete
24.5
Gpa
Poisson’s ration of shotcrete
0.3
Density of shotcrete
2200
kg/m3
Rayleigh damping is used in the seismic analysis of soil nailwalls. Rayleigh damping is specified in FLAC with theparameters fmin in Hertz (cycles per second) and ξ min . For
geological materials, damping (ξ min ) commonly falls in therange of 2 to 5% of critical; for structural systems, 2 to 10% isrepresentative. [3]The idea in the dynamic analysis is to adjust f min of theRayleigh damping so that it is 3:1 range coincides with therange of predominant frequencies in the problem. The“predominant frequencies” are neither the input frequencies
nor the natural modes of the system, but a combination of
both. The idea is to try to get the right damping for theimportant frequencies in the problem. [4]Numerical distortion of the propagating wave can occur in a
dynamic analysis as a function of the modeling conditions.
Both the frequency content of the input wave and thewave-speed characteristics of the system will affect thenumerical accuracy of wave transmission. Kuhlemeyer andLysmer (1973) show that for accurate representation of wave
transmission through a model, the spatial element size, l, mustbe smaller than approximately one-tenth to one-eighth of the
wavelength associated with the highest frequency componentof the input wave [5]. In this research, finite element meshsize is considered for accuracy of wave transmission.
4. Influence of nail length in static and dynamic analysis

499

The numerical model that is described in part 3 is used toinvestigate the influence of nail length during static anddynamic analysis. The only exception is the variation in naillength that is varied in 5, 7, 9 and 11 meters. The dynamic
analysis is carried out by Manjil velocity history. Fig. 5 shows
lateral displacement of the wall during static and dynamic
analysis. As shown in Fig. 5 displacement behaviour defers inthe static and dynamic analysis. The maximum displacementof the wall in the dynamic analysis happens on the top of thewall. Other than the displacement of the wall in the static
analysis happens on the one-third height of the wall. The
variation of nail length effects on the stability of the wall inthe dynamic analysis more than static stability of the wall.
Fig. 6 shows maximum normalized axial force of the nails
during static and dynamic analysis. As shown in Fig. 6
mobilized axial force behaviour nearly is same in the staticand dynamic analysis.

analysis. As shown in Fig. 7 displacement behaviour defers inthe static and dynamic analysis. The maximum displacementof the wall in the dynamic analysis happens on the top of thewall. Other than the displacement of the wall in the static
analysis happens on the one-third height of the wall. As
shown in Fig. 7 minimum displacement of the wall occurs in
the models with 10 and 20 degrees of inclination angle.
Otherwise walls with 0 and 30 degrees of inclination angle
have larger displacement. This situation shows that there is an
optimum inclination angle in both static and dynamic
manners between 10 and 20 degrees.
Fig. 8 shows maximum normalized axial force of the nails
during static and dynamic analysis. As shown in Fig. 8 there
is similar function of nail axial force in both dynamic andstatic analysis. In the upper nails, lower inclination of nails
causes lower axial force. Otherwise in the downer nails, lower
inclination of nails causes higher axial force.

Fig. 5) Lateral displacement of the wall in static and dynamic
analysisFig. 7) lateral displacement of the wall during static anddynamic analysis

Fig. 8) Maximum normalized axial force of the nails during
static and dynamic analysis.

Fig. 6) Maximum normalized axial force of the wall in staticand dynamic analysis
6. Influence of nail inclination angle in static and dynamic
analysis
The same numerical model is used to investigate the influenceof nail inclination angle during static and dynamic analysis.
The only exception is the variation in nail inclination angle
that is varied in 0, 10, 20 and 30 degrees. The dynamic
analysis is carried out by Manjil velocity history. Fig. 7 shows
lateral displacement of the wall during static and dynamic

7. Conclusion
In this paper static and dynamic analysis is carried out on asoil nail model. Manjil earthquake acceleration is selected forthe dynamic analysis. Two main geometry parameters of soil
nail walls are considered for comparison between dynamic
and static behaviour of soil nail walls: nail length and nailinclination angle. The results are gained by lateraldisplacement of the wall and maximum axial nail force.

500

These two parameters can explain the external and internal
response of the system respectively.
The results show that the maximum displacement of the wallin the dynamic analysis happens on the top of the wall. Other
than the displacement of the wall in the static analysis
happens on the one-third height of the wall. The variation ofnail length effects on the stability of the wall in the dynamic
analysis more than static stability of the wall. In the upper
nails, lower inclination of nails causes lower axial force.
Otherwise in the downer nails, lower inclination of nails
causes higher axial force.
Reference[1] Thompson, S. R., and Miller, I. R ., “Design, Construction
and Performance of a Soil Nailed wall in Seattle,
Washington”, Design and Performance of Earth Retaining
Structures, Geotechnical Special publication, No. 25, ASCE,
1990, pp. 629-643.
[2] W.B. Wei, Y.M. Cheng, “Soil nailed slope by strength
reduction and limit equilibrium methods”, Computers andGeotechnics journal, 2010.
[3] Biggs, J. M. Introduction to Structural Dynamics. New
York: McGraw-Hill, 1964.
[4] Itasca Consulting Group, Inc. , “Fast Lagrangian Analysis
of Continua in 2 Dimensions, FLAC2D” Dynamic Analysis,
2005, pp 13[5] Kuhlemeyer, R. L., and J. Lysmer. “Finite Element
Method Accuracy forWave Propagation Problems,” J. Soil
Mech. & Foundations, Div. ASCE, 1973, 99(SM5), 421-427.

501

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia,Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

SHEAR STRENGTH IMPROVEMENT OF FIBROUS PEAT DUE TO
CONSOLIDATION PRESSURE
Vivi Anggraini1, Nurly Gofar2 and Bujang bin Kim Huat1
Universiti Putra Malaysia,2Universiti Teknologi Malaysia, Department of Civil Engineering, Malaysia

1

ABSTRACT
A major problem related to construction on peat deposit is the low shear strength of the soil. However previous researches
indicated that the shear strength could increase upon consolidation and hence some improvement methods have beendeveloped to increase the bearing capacity of the peat deposit by utilizing consolidation pressure. The aim of this paper isto evaluate the shear strength improvement of fibrous peat due to consolidation. The initial undrained shear strength was
obtained from vane shear test in-situ. Subsequently, samples were prepared under several consolidation pressures: 50
kPa, 100 kPa, 150 kPa, and 200 kPa. The shear strength of the samples was obtained by triaxial compression test under
unconsolidated undrained condition. The results showed that there is an increase in shear strength due to application of
consolidation pressure as compared to the initial shear strength obtained by field vane shear test. However, the ratio ofstrength increase for peat in this study is only 0.06 which is much lower than previously published data may be due tohigh fiber content.
Keywords: Peat Soil, Shear Strength, Unconsolidated Undrained, Consolidation Pressure
1. INTRODUCTION
Initial stability of construction on peat is the most criticalproblem due to low undrained shear strength of peat innormally consolidated state and hence low bearing capacityof foundation soil. Conventional solution is to replace thepoor soil by suitable imported fill materials. This practice isnaturally very expensive. In addition, there must be anenvironmentally acceptable location to waste excavated soil
within an economically acceptable haul distance [1]. Thismethod also need maintenance work related to long term
settlement and horizontal movement [2]. The fill induces aconsolidation pressure, thus increases the shear strength ofthe peat. Laboratory compression tests have to be performed
in order to estimate and take into account the rate of increaseof undrained shear strength as a function of consolidation andsubsequent densification of peat. The undrained shear
strength of peat is typically determined by vane shear in thefield and undrained triaxial compression test in thelaboratory. The ratio of increase can be applied to the in-situ
shear strength obtained from vane shear test. Therefore thethickness of fill for each stage can be increased as aconsequence of the increase in shear strength.
Based on his study, Magnan [2] suggested a ratio of shear
strength increase due to increase in overburden pressure of
0.5 for peat soil. Furthermore, Edil and Wang [3] collected
normalized undrained strength (c u /σ’ 3c or c u /σ’ 1c ) as afunction of organic content for all peat and organic soil. The
study showed that fibrous and amorphous peat presents noperceptible differences and gives an average

normalized undrained strength to consolidation pressure of
0.59 with most of the data falling between 0.5 and 0.7. Thesec u /σ’ v values appear unusually high compared to the typicalvalues for inorganic clays that lay between 0.2 and 0.25. Chen
and Tan [4] found that the ratio of shear strength increase due
to consolidation for clay obtained from Klang, Malaysia is0.25. However, there is no published data for the ratio ofshear strength increase for peat found in Malaysia.
2. EXPERIMENTAL PROGRAM
2.1 Physical and Chemical Properties
The peat sample for this study was obtained from Pontian,Johor. Observation made in the location showed that the peat
is categorized as deep peat with thickness of more than 5 m.
Ground water table exists at depth less than 1 m at the time ofsampling, thus samples were taken from depth of 1 to 2 mfrom ground surface. The samples were taken below watertable; thus it is expected to be in fully saturated condition.
Visual identification showed that the peat is dark brown, very
soft, and contains a large amount of fiber. Plant structures
such as roots are easily recognizable from the soil. Long,
slender roots, and rootlets are identified as the remaining offorest vegetation. The texture is coarse and may results inlarge permeability.
In-situ measurement of water content was not possible. Thus,sufficient care was taken during the sampling of the peat inorder to maintain the natural water content. The preliminary
identification of the soil was made based on the index
properties and classification tests. Index properties include
the determination of water content, specific gravity, bulk unit

502

weight, and the initial void ratio. The summary of index
properties is presented in Table 1.
Table 1.The summary of index properties of peat soil in West
Malaysia peat [5]Parameters

Results

Natural moisture content (%)
Specific Gravity (Gs)
Bulk unit weight (kN/m3)
Dry unit weight(kN/m3)
Initial void ratio (e 0 )Acidity (pH)

608
1.47
10.02
1.4
8.92
3.24

Published data
(ranges)200-700[6]
1.30-1.90[6]
8.30-11.50[6]
1.00-1.65[7]
3.00-15[6]
3.00-4.5[8]

The ash content is 3 %. The fiber within the range of
published data on peat soil content of 90 % is considered very
high as compared to published data around the world (Table2) but this is a typical fibrous peat obtained in West coast ofPeninsular Malaysia [6],[8]. The average percentage ofparticle passing from 0.063 mm sieve is 2.37 % which showthat the soil contain a large amount of fiber. The summary of
the classification tests results is presented in Table 2.
Table 2.The summary classification test results in West
Malaysia peat [5]Parameters

The average natural water content obtained from laboratory
tests is 608 %which indicates that the peat has a highwater-holding capacity. This value is within the range
obtained by previous researches for peat soil in West
Malaysia. The average specific gravity obtained usingkerosene on pycnometer test is 1.47 and it is within the range
for fibrous peat. The initial void ratio is 8.92 which is within
the range given by Huat [6]. The average unit weight of thepeat is 10.02 kN/m3 which give a bulk density of 1.002
Mg/m3 which is also within the range given in Table 1. The
dry unit weight of the peat is 1.40 kN/m3 and it is slightly lessthan predicted by Al-Raziqi et al. [7]. From the information
indicated above, the unit weight of the peat in this study isclose to the unit weight of water; hence in-situ effective stressis very low and the void ratio of the peat is very large.
The void ratio also includes the volume of gas generated
during decomposition process. The average void ratio for the
51 fibrous peat obtained in Pontian is 8.92 and this is within
the range given for West Malaysian peat (3 -15) presented by
Huat [6]. Peat in Malaysian Peninsular is known to have low
pH value and the acidity tends to decrease with depths. The
test results showed that the average pH value of the fibrouspeat used in this study is 3.24 which is in the lower side of therangepublished for Malaysia peat (3.0 - 4.5) presented by
Muttalib et al. [8].
2.2. Classification
The peat in this study was classified based on the degree of
humification [9] and the organic and the fiber content. The
von Post scale is based on the appearance of soil water that is
extruded when a sample of the soil is squeezed in the hand
and the soil is classified based on Table 2. When the test was
performed on the samples used in this study, brown water
comes out from the soil and the soil left in hand has a large
amount of fiber; thus the peat can be classified as fibrous peatwith H4 degree of decomposition according to von Post scale.
The organic content the peat is 97 % which is quite high but
within the range of published data on peat soil found in
Malaysia.

Von Post humificationof peatOrganic content (%)
Ash content (%)
Fiber content (%)

Results
H4

Published data
(ranges)
H1-H4
[9]

97
3
90

> 90
< 10
> 20

[6]
[6]
[10]

2.3 In-situ Shear Strength
The in-situ shear strength was obtained by field vane test
conducted at depth of 1 and 2 m in the site. The average initial
undrained shear strength of peat obtained by the field vaneshear test is 10.10 kPa which is comparable to the range of theundrained shear strength of peat in West Malaysia (3-15 kPa)
and Sarawak (10 kPa) as indicated by Huat [6]. The peat is
identified as very sensitive to disturbance with sensitivity of
5.64, which is also within the range of published data 4-12
[7].
2.4 Remoulded Sample Preparation
Remolded sampler preparation equipment was used for soil
sample preparation. The remolded sampler was designed asshown in Fig 1. Soil sample was put into the remolded
sampler equipment. The plate with some holes was placed at
the bottom of the sampler to drain water from the sample. The
soil sample was then loaded using steel load platen. Pressure
was applied on the steel load platen by compressed air. The
air pressure supply was from the centralized air compressor
and pressure panel system in the geotechnical engineeringlaboratory, Universiti Teknologi Malaysia (UTM). Air
pressure applied to the soil sample was based on the
maximum applied pressure needed for remolded sample
preparation. The pressure for the remolded sample wasadjusted from the dial gauge attached on the top of the load
platen. Samples were placed into 152 mm diameter remolded
sampler equipment under various consolidation pressure;consist of 50 kPa, 100 kPa, 150 kPa and 200kPa. The
consolidation was assumed completed when there is no morewater draining out from the bottom of the remolded sample
equipment.

503

Fig.2 shows the relationship between principal stressdifference (σ 1 -σ 3 ) and the axial strain from triaxial test under
UU condition and cell pressures 50 kPa, 100 kPa and 150 kPa.The results show the increase in shear strength up to a peakvalue and then the reduction in shear strength. The maximum
deviator stress for sample A1 with cell pressure of 50 kPa is26.7 kPa, for sample B1 with cell pressure 100 kPa is 28.10
kPa and for sample C1 with cell pressure 150 kPa is 40.6 kPa.
Axial strain at maximum deviator stress is less than 5 %which shows that the sample is of high quality.

Mohr Circle

Shear stress (kPa)

60

Fig. 1: Remoulded sampler preparation equipment
3. RESULT AND DISCUSSION

Deviator stress σ1-σ3 (kPa)

Test C1
Test B1

Test A1

50 kPa100 kPa
150 kPa

1

2

3

4

0

20

40

60

80

100

120

140

160

180

Fig.3: Shear stress at failure (τ f ) versus normal stress (σ n )Twelve sets of test containing three soil samples wereconducted on Triaxial test apparatus under Unconsolidated
Undrained (UU) condition [11]. In this project the tests werecarried out in the triaxial apparatus on the specimens ofdiameter 38 mm and height 76 mm and were subjected to cellpressures of 50 kPa, 100 kPa, 150 kPa, and 200 kPa. The
results obtained from the triaxial test were used to analyze theshear strength parameters of fibrous peat soil from Pontian bydetermining the undrained cohesion (c u ) and the angle ofinternal friction (φ u ). The test results were used to evaluate
the increase of shear strength due to consolidation pressure.

020Normal stress (kPa)

3.1 Shear Strength Obtained from Triaxial Test

45
40
35
30
25
20
15
10
5
0

Set1
Set 1
ccu
kPa
u == 9.23
9.23 kPa
o
φφuu=
=3.87
3.87°

40

5

6

7

8

Axial strain ε (%)
Fig. 2:Typical curve for deviator stress versus axial strain for
set 1

Fig.3 shows the typical Mohr circle obtained from set 1
sample for each specimen. The shear stress at failure (τ f ) isplotted against the corresponding normal stress (σ n ). A linethat the best fit through the corresponding points of the graphis drawn and results in the cohesion of 9.23 kPa and the angle
of internal friction (φ u ) of the failure envelope is 3.87o. The
angle of internal friction obtained from the test is very close to0.
The angle of internal friction (φ u ) for undrained cohesion istheoretically equal to zero but the range of 0o-5o is acceptable
due to the presence of air in the specimen resulted in perfectundrained condition is maintained during the test [12].
Shearing
inducedbyunconfinedcompression,
unconsolidated undrained compression and consolidated
undrained compression did not result in the formation of aplane of failure. The shear strength obtained from unconfinedcompression tests increased from a negligible value to about
25 kPa, while the envelope of Mohr circles of total stress forconsolidated undrained tests was inclined at an angle of 5o[13]. Air being highly compressible and reduces air voids
proportionally with higher lateral pressures, thereby reduce
the total volume of the specimen under testing.
Table.3 shows the summary of triaxial compression test
results for each test set under unconsolidated undrained
condition in terms of undrained cohesion and undrained angle
of internal friction. The average cohesion is 10.44 kPa and theaverage angle of internal friction is 3.60o.

504

200

Table 3. Result of Triaxial Compression Test

Set 1
Set 2
Set 3
Set 4
Set 5
Set 6
Average

Parameter of Shear Strength
Cohesion c u (kPa)

Angle of friction φ u o

9.23
6.36
10.84
12.5
15.74
8.98
10.44

3.87
4.11
3.91
3.21
2.2
4.34
3.6

This relationship is illustrated in Fig 4. It can be observed that
there is not much scattered of data obtained from the test
indicated by R squared value of 0.93 which gives a strong
relationship between undrained shear strength andconsolidation pressure.
Undrained shear stress cu (kPa)Testc u(initial) = Initial shear strength obtained from vane shear test,= Ratio of strength increase, and= Consolidation pressure.

λσ vc

3.2 Effect Consolidation on Strength
The effect of consolidation pressure on the shear strength ofpeat is studied by performing triaxial compression test under
UU condition on samples prepared under different
consolidation pressure. In this study, consolidation pressureof 50 kPa, 100kPa, 150 kPa and 200 kPa were used. The
results are shown in Table 4. The average undrained shear
strength from test results from four different consolidationpressures are 13 kPa, 15.66 kPa, 21 kPa and 21.66 kPa.
Table 4.Result summary of shear strength underconsolidation pressure
Consolidation
Initialcu
Pressure
Samplecellcu
Average(kPa)pressure
(kPa)
A1
11
50
A2
50
13
13
A3
15
B1
11
100
B2
100
16
15.66
B3
20
C1
19
150
C2
150
21
21
C3
23
D1
19
200
D2
200
21
21.66
D3
25
It can be seen from Table 4 that the undrained cohesion (c u )increases as the consolidation pressure increases. These data,
together with the in-situ shear strength obtained by vane shear
test were used to develop the correlation between the shear
strength and the consolidation pressure. The relationshipbetween the strength of peat under consolidation pressure andinitial shear strength and the effective consolidation pressure
can be expressed in the following form:c u(mod) = c u(initial) +λσ vc
where:c u(mod) = Shear strength under consolidation pressure,

(1)

50
40
y = 0.06x + 10.1030
20

R² = 0.93

10
0

0

50

100

150

200

250

Vertical consolidation pressure σvc (kPa)Fig 4. Undrained shear stress c u versus vertical consolidationpressure σ vc
The equation relating the shear strength and the effective
consolidation pressure for fibrous peat used in this study is:
C u(mod) = 10.10 +0.06 σ vc

(2)

It can be seen from Eq.(1) that the initial undrained shear
strength of the peat soil predicted by the equation agrees withaverage undrained shear strength obtained by field vane shear
test. The initial value of undrained shear strength from vaneshear in the field (c u mod ) is 10.10 kPa. The increase inundrained shear strength is small, however, very small ascompared to the published data may be due to the anisotropyof the fiber in the soil. The relationship shown in Eq.(2)allows the estimation of strength allows a means of estimation
strength gain due to consolidation pressure and to get themodified shear strength c u (mod) of the peat. The result provesthat the shear strength value can be improved when subjected
to consolidation pressure; eventhough the increase is small.
4. SUMMARY AND CONCLUSION
The study on the shear strength improvement of peat due toapplication of consolidation pressure was conducted on thisproject for fibrous peat sampled from Pontian, Johor. The
objective was to find the ratio of strength increase with theincrease in consolidation pressure. Consolidation pressure
applied to the soil prior to Triaxial Compression Test under
Undrained Unconsolidated condition have the effect ofincreasing the shear strength obtained from the test. The

505

average value of undrained shear strength from test results are
13 kPa, 15.66 kPa, 21 kPa and 21.66 kPa under consolidationpressure of 50 kPa, 100kPa, 150 kPa, and 200 kPa
respectively. The relationship derived from the Triaxial test
results shows an initial undrained shear strength equal to thatobtained from field vane shear test (c u = 10.10 kPa).
However, the ratio of strength increase is only 0.06 which ismuch lower compared to previously published data may be
due to high fiber content.
NOTATION
The following symbols are used in this paper:cc’cu
c u(mod)c u(initial)

φ
φu
τf
τ
ε
σn
σ’ v
σv c
σ1
σ2
σ3
σ’ 1c
σ’ 3c
σ 1 -σ 3
σ 1 ’- σ 3 ’
σ 1 /σ 3
σ 1 ’/σ 3 ’λ cu
λCohesion value of soil
Apparent cohesion in terms of effective stress
Undrained shear stress
Undrained shear stress modify
Undrained shear stress initial
Angle of internal friction soil
Undrained angle of internal friction soil
Shear stress at failure of soil
Shear stress of soil
StrainNormal stress due to applied vertical load
Effective vertical consolidation pressure
Vertical consolidation pressure
Major principal stresses
Intermediate principal stresses
Minor principal stresses
Effective major principal stresses
Effective minor principal stresses
Principal total stress difference
Principal effective stress difference
Principal effective stress ratio
Principal total stress ratio
Coefficient ratio of undrained shear strength
Coefficient ratio of shear strength

REFERENCES[2] Magnan, J (1994). Construction on peat. State of the art
in France. Advances Understanding and Modelling themechanical behaviour of peat, ed. Den Haan etal.,1994.pp.369-380.
[3] Edil, T. B. and Dhowian, A. W. At-rest Lateral Pressureof Peat Soils. Conf. on Sedimentation and Consolidation
Model, ASCE, San Fransisco.1981.pp 411-424.
[4] Chen, C.S & S.M Tan. Some engineering properties ofsoft clay from Klang Area.2nd Int.Conference on
Advances
in
Soft
Soil
EngineeringandTechnology.2003. pp.79-87, 2-4 July 2004, Putra Jaya.
[5] Nurly Gofar and Yulindasari Sutejo. Properties of
Fibrous Peat. Senai (Malaysia): Seminar Penyelidikan
Kejuruteraan Awam (SEPKA), Universiti Teknologi
Malaysia (UTM).2005.
[6] Huat, B. B. K.. Organic and Peat soil engineering. Univ
Putra Malaysia Press.2004.ch.3 and ch.4.
[7] Al-Raziqi, A. A., Huat, B. B. K. and Munzir, H. A.
Potential Usage of Hyperbolic Method for Prediction ofOrganics Soil Settlement. In Proceeding of 2ndInternational Conferences on Advances in Soft Soil
Engineering and Technology, ed. Huat et al., Putrajaya
Malaysia.2003. pp 439-445.
[8] Mutallib,
A.A.,J.S.
Lim,
M.H.
Wong
&
L.Koonvai.Characterisation, distribution and utilizationof peat in Malaysia. In Proceedings International
Symposium on Tropical Peat Land, 2001. pp.7-16,
Kuching, Sarawak.
[9] von Post, L. Sveriges Geologiska Undersoknings
Torvinventering Och nagre av Dess Hittills Vunna
Resultat, Sr.Mosskulturfor. Tidskr 1.1992. pp.1-27.
[10] Molenkamp,F. Investigation of requirements for plavestrain elements tests on peat. In Advances inUnderstanding and Modeling the Mechanical behaviourof Peat, ed. Den haan et al. Balkema.1994.
[11] Head, K.H. Manual of Soil laboratory Testing, Volume
3. Pentech Press, London.1981.
[12] Mandal J. N. Soil Testing in Civil Engineering.A.A
Balkema. Rotterdam.1995.
[13] Hanrahan, E. T. An Investigation of Some Physical
Properties of Peat. Geotechnique, London, England,4(2).1954.pp.108-123.

[1] Jarrett, P.M. Geoguide 6. Site Investigation for organicsoils and peat. JKR: Document 20709-0341-95. Institut
Kerja Raya Malaysia.1995.

506

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Factors affecting the stability analysis of earth dam slopes subjected toreservoir drawdown
T. Souliyavong1, C. Gallage1, P. Egodawatta1 and B. Maher2
Queensland University of Technology, Australia. 2South East Queensland Water, Australia

1

ABSTRACTThe drawdown of reservoirs can significantly affect the stability of upstream slopes of earth dams. This is due to theremoval of the balancing hydraulic forces acting on the dams and the undrained condition within the upstream slope
soils. In such scenarios, the stability of the slopes can be influenced by a range of factors including drawdown rates,
slope inclination and soil properties. This paper investigates the effects of drawdown rate, saturated hydraulicconductivity and unsaturated shear strength of dam materials on the stability of the upstream slope of an earth dam. In
this study, the analysis of pore-water pressure changes within the upstream slope during reservoir drawdown was
coupled with the slope stability analysis using the general limit equilibrium method. The results of the analysis
suggested that a decrease in the reservoir water level caused the stability of the upstream slope to decrease. The dam
embankment constructed with highly permeable soil was found to be more stable during drawdown scenarios,
compared to others. Further, lower drawdown rates resulted in a higher safety factor for the upstream slope. Also, thesafety factor of the slope calculated using saturated shear strength properties of the dam materials was slightly higher
than that calculated using unsaturated shear strength properties. In general, for all the scenarios analysed, the lowest
safety factor was found to be at the reservoir water level of about 2/3 of drawdown regime.
Keywords: reservoir drawdown, earth dam, slope stability, hydraulic conductivity, unsaturated shear strength.

1. INTRODUCTION
Rapid drawdown of storage water level is a critical design
mechanism for most reservoirs particularly in situations suchas post earthquake damage or piping events. Rapid
drawdown of water levels in reservoirs with earth dams canlead to failure of the upstream slope. A number of such
failures are listed by Morgenstern [8].
Prior to reservoir drawdown, steady-state seepage conditions
can be assumed within the dam materials and the hydrostaticpressure resulting from the reservoir water helps to partiallystabilize the upstream dam slope. The reservoir drawdown
could establish undrained conditions and development ofexcess pore-water pressures that could dissipate with time(transient seepage). These scenarios could cause instability
and consequently lead to failure of the slope. The instabilityof the slope during a reservoir drawdown can be affected bya range of factors including drawdown rates, slope
inclinations and soil properties of earth dam materials [1],[2], [7], [8], [11]. Transient seepage and the instability of theearth dam slope induced by rapid drawdown can be affectedby the unsaturated properties of dam materials such as
hydraulic conductivity functions and unsaturated shear
strength.
This paper presents the analytical outcomes of an earth dam
during a drawdown regime. The analysis undertaken

considered the effects of the saturated and unsaturated
properties of dam materials, the saturated hydraulicconductivity of the dam embankment soil, and thedrawdown rate of the reservoir on the upstream slope
stability of the dam. The influences of pore-water pressuresor transient seepage on the slope stability were analysed bycoupling SEEP/W [5] with a stability analysis tool,
SLOPE/W [6]. The outcomes generated from this analysis
could form the basis for determining a safe drawdown rate
and a safe drawdown level for some Queensland earth dams.
2. MODEL DAM AND MATERIAL PROPERTIESThe dam configuration shown in Fig. 1 and material
properties listed in Table 1 were used in this study. The
dimensions, material zones and soil properties of the dam
cross section were chosen to be generic representatives ofexisting earth dams in Queensland, Australia. For this,geometries and material properties of a range of earth dams
were studied based on the data provided by the Queensland
Bulk Water Supply Authority (Seqwater); the dam owner inSouth East Queensland.
Apart from basic geometry and material properties, the soil
water characteristic curves (SWCCs) form a critical part in
the analysis of the earth dam with unsaturated material
conditions. Since the measured SWCCs for the dam
materials were not available, SWCCs (see in Fig. 2) assigned
for the clay core, embankment soil and sand filter were
estimated from their grain-size data [4].

507

5.3m
16.2m

Embankment

19°
9.2m

Sand filter/drainCore

H=25.4m

19°Foundation
Bedrock

30m

111.6m30m

Fig. 1. Geometry and material zones of the earth dam
Table 1. Properties of the earth dam materials
Material properties

Volumetric water content (m3/m3)

Saturated hydraulic conductivity
Saturated unit weight
Effective cohesion
Effective friction angle
Volume compressibility coefficient
Volumetric water content
Diameter at 10% passing on
a grain–size curve
Diameter at 30% passing on
a grain–size curve
Diameter at 60% passing on
a grain–size curve
Liquid limit
Plastic limit
0.5

m/sec
kN/m3
kPa
degreem2/kN
m3/m3

Clay core
1.2x10-1020729
18x10-5
0.4

Value
Embankment Sand filter
3x10-74.9x10-3
20.5
20.5
18
0
30.8
34
12 x10-6
0
0.35
0.31

Foundation
3.9x10-6
19.2
21
32.3
12x10-6
0.33

D 10

mm

0

0.06

0.3

0.05

D 30

mm

0.004

0.6

1

0.5

D 60

mm

0.07

0.8

1.6

1.2

LL
PL

percent
percent

51
20

30
16

0
0

31
13

Symbol

Unit

k sat
γ sat
c’
φ’
Mv
θw

3.1 Seepage Analysis
The finite element (FE) method available in SEEP/W was
employed to simulate 2-D steady-state and transient seepage
in the earth dam before and during the rapid drawdown,
respectively. This program solves the general governing
differential equation for two-dimensional seepage shown in
Eq. (1) using FE method.

Embankment
Clay core
Sand filter

0.4
0.3
0.2

∂ 
∂H 
∂ 
∂H 
∂θ
kx
+Q =
+
kx
∂y 
∂x 
∂x  ∂y 
∂t

0.1
0

0.1

1

10

100

1000

10000

(1)

where H = total head, k x = hydraulic conductivity in the xdirection; k y = hydraulic conductivity in y-direction; Q =applied boundary flux; θ = volumetric water content; and t =time.

100000Matric suction (kPa)
Fig. 2. SWCCs for dam materials
3. MODELLING AND ANALYSISThe analysis undertaken includes transient seepage analysis
to determine pore-water pressures during drawdown andstability analysis of the upstream dam slope. Both analyses
were conducted in automatic coupled mode. For the analysis,
each material zone was assumed to be homogeneous andisotropic.

The FE mesh shown in Fig. 3 was used to replicate the dam
structure. The mesh was defined as an unstructured patternof quadrilateral and triangular elements with 3 and 4 nodes
respectively. Prior to drawdown simulation, the dam wasanalysed for steady-state seepage conditions, assuming thereservoir water level at the dam crest (elevation of 25.4 m) isa critical situation. For this, the boundaries ABC and DE in
Fig. 3 were defined as constant total head boundaries withthe values of 25.4 m and 9.2 m respectively. The
corresponding saturated hydraulic conductivity values508

tabulated in Table 1 were assigned for the defined material
zones in the dam.

Since the drawdown of reservoir produces unsaturated
conditions in the dam materials such as clay core,embankment and sand filter, the unsaturated shear strengthsof the materials were taken into account. In this analysis, Eq.
(2) proposed by Vanapalli et al. [10] was used to estimate
the unsaturated shear strength of the materials. Therein,
effective shear strength parameters and SWCCs were needed
to estimate the unsaturated shear strength of the materials.

Outcomes of the steady-state seepage analysis were thenused to define the initial pore-water pressure distribution for
the transient seepage analysis. To simulate the reservoir
drawdown numerically, the upstream boundary (ABC)condition was defined by a total head function. This type ofboundary will allow the user to define the change in totalhead as a function of time. For transient seepage analysis,
the total head of the upstream boundary was decreased
linearly from 25.4 m to 9.2 m at different rates of thereservoir drawdown. During the drawdown, the conditionsof the dam materials could change from saturated tounsaturated and thereafter the saturated hydraulicconductivity of each dam material should be defined as afunction of suction. Since the laboratory measured hydraulicconductivity functions of the dam materials were notavailable, Fredlund and Xing’s method [3] was used toestimate the hydraulic conductivity function of each materialusing its SWCC and saturated hydraulic conductivity. The
transient seepage analysis of the dam during reservoir
drawdown will give the variation of pore-water pressureswith drawdown time throughout the FE mesh.

 θ − θ r  tan ϕ '  (2)τ = c ' + (σ n − u a ) tan ϕ ' + (u a − u w )  w
 θ s − θ r where τ = shear strength of saturated/unsaturated soil; c’=effective cohesion of unsaturated soil; σn = total normal
stress on the failure plane of the soil; φ’ = effective friction
angle; (ua– uw) = matric suction in the failure plane of thesoil; ua = pore-air pressure in unsaturated soil; uw = porewater pressure; θw = volumetric water content; θs = saturated
volumetric water content; and θr = residual volumetric watercontent which can be assumed to be equal to 10% of θs.
In this study, pore-water pressure changes obtained from thetransient seepage analysis were used for the stability analysisof the upstream dam slope at different stages of drawdown.

3.2 Stability Analysis
To analyse the stability of upstream slope of the dam during
drawdown, the GLE method available in SLOPE/W was
employed. In this study, the entry (DEFG) and exit (ABC)method was used to define trial failure surfaces (as shown in
Fig. 4) for each simulation scenario. The software analyses
all the trial surfaces to obtain a critical surface with aminimum safety factor for each drawdown scenario.
30

4.1 Effects of Saturated Hydraulic Conductivity of
Embankment Soil on the Slope Stability
To investigate the effects of saturated hydraulic conductivity(k sat ) of embankment soil on the stability of the upstream
dam slope in terms of safety factor (FOS), the analysis
C

25

Elevation (m)4. RESULTS AND DISCUSSIONS

Drawdown

20
15

D

B

A

10

E

5
0

0

5

10 15 20 25 30 35 40 45 50 55 60 65 70 75 80 85 90 95 100 105 110 115 120 125 130 135 140 145 150 155 160 165 170 175

Distance (m)Fig. 3. Finite element mesh of the earth dam used in SEEP/W
3.915

30

D

Elevation (m)

25
20

E

F

G

C

15

A

10

B

5
0

0

5

10 15 20 25 30 35 40 45 50 55 60 65 70 75 80 85 90 95 100 105 110 115 120 125 130 135 140 145 150 155 160 165 170 175

Distance (m)Fig. 4. Model dam and some of trial failure surfaces used in stability analysis in SLOPE/W
509

4.6

ksat=3x10-3 m/s
ksat=3x10-5 m/s
ksat=3x10-7 m/s
ksat=3x10-9 m/s

4.2

FOS

3.8
3.4
3
2.6
2.2
1.8

Drawdown 1 m/day
26

24

22

20

18

16

14

Reservoir drawdown (m)

12

10

8

Fig. 5. Variation of FOS of upstream dam slope withreservoir drawdown and unsaturated soils.
As shown in Fig. 5, for a given k sat of embankment soil, the
FOS of the upstream slope of the earth dam decreases withthe reservoir drawdown and reaches the lowest value when
the reservoir is lowered about 1/2 to 2/3 of the adopted
drawdown regime (16.2 m). The removal of the balancing
hydraulic forces acting on the upstream slope due to thereservoir drawdown is the main reason for a decrease in the
FOS of the slope. The temporary undrained condition within
the dam created by the drawdown leads to a decrease in theshear strengths of the dam materials which, in turn, cause the
FOS of the slope to decrease. As shown in Eq. (2), an
increase in pore-water pressures causes a decrease in theshear strengths of the dam materials. The FOS of the slope
increases slowly from its minimum value as the reservoir
water is further drawn. When the reservoir level is reduced
from 1/2 to 2/3 to the end of the drawdown regime, the shear
strengths of the dam materials increase due to the pore-water
dissipation. This has positive effects on the stability of theslope.
Saturated hydraulic conductivity (k sat ) of the embankment
soil exerts a significant influence on the upstream slope
stability during reservoir drawdown. It can be seen from Fig.
5 that during the early stages of drawdown (0 to 1/4 ofdrawdown regime) the upstream slope of the dam
embankment constructed with low k sat soil is more stable
than that constructed with high k sat soil. However, as thereservoir level decreases further, this behaviour changes bymaking the high k sat embankment soil more stable than the
low k sat embankment soil. It was noted that the saturated
zone (area below the phreatic surface) under the steady–state

seepage conditions is larger for the high k sat embankment
soil than the low k sat soil. Since a high k sat embankment soil
allows water to penetrate deeper into the dam, the productionof a larger saturated zone would be the primary reason forhaving the low FOS during the early stages of drawdown forthe high k sat embankment soil. The pore-water pressures in asmaller saturated area within the low k sat soil may dissipate
faster than that of a larger saturated area within the high k sat
soil and so an embankment slope with less excess pore-waterpressures is more stable than that with more excess porewater pressure or a larger saturated area. When the reservoir
water level is decreased further (about 1/4 to the end of thedrawdown regime), the difference between the areas of thesaturated zone within the dam constructed with the low k sat
embankment soil and that within the high k sat soil is of noconsequence. Also, the pore-water pressure within the highk sat embankment soil can dissipate faster than that within the
low k sat soil. As a result, at the given drawdown level, theshear strength of the high k sat soil is larger than that of thelow k sat soil.
In this study, the effects of using saturated and unsaturated
shear strength properties for the stability of the upstream
slope during rapid drawdown of the reservoir were
investigated. The pore-water pressures obtained from thetransient seepage analyses conducted at the reservoir
drawdown rate of 1m/day for different saturated hydraulicconductivity values of the embankment soil (e.g., 3x10-3 and3x10-7 m/s) were used for stability analysis with saturated
shear strength (τ sat ) and unsaturated shear strength (τ unsat )properties of dam materials. Hydraulic conductivity
functions and SWCCs were not considered for the
calculations of saturated shear strengths of the dam
materials.
4.2

Ksat=3x10-3 m/s (τunsat)
Ksat=3x10-3 m/s (τsat)
Ksat=3x10-7 m/s (τunsat)
Ksat=3x10-7 m/s (τsat)

3.8
3.4

FOS

described in section 3 was repeated for different k sat valuesof embankment soil (e.g., 3x10-3, 3x10-5, 3x10-7, 3x10-9m/s). For this, the reservoir drawdown rate of 1 m/day wasused. It is reported that the common range of drawdown rateis from 0.5 to 1 m/day in dam engineering applications [9].
The SWCCs of the dam materials were used for the entire
analyses. The results of the analyses are shown in Fig. 5.

3

2.6
2.2

Drawdown 1 m/day
26

24

22

20

18

16

14

12

10

8

Reservoir drawdown (m)Fig. 6. Variation of FOS of upstream dam slope with
unsaturated and saturated soils.
The results shown in Fig. 6 depict that the trend of the FOS
variation of the upstream dam slope with saturated shear
strength is similar to that observed for the dam materials
with unsaturated shear strength. During the drawdown, the510

slope stability calculated using saturated shear strength isslightly higher than that calculated using unsaturated shear
strength. It is noted that the saturated (wetting) zone under
the steady–state seepage is larger in the dam with
unsaturated materials than that with saturated soils. Since theunsaturated materials allow water to penetrate deeper into
the dam due to soil suction, the production of a larger
wetting zone would be a primary reason for having the low
FOS for the dam with unsaturated materials. The pore-waterpressures in a smaller wetting zone within the dam withsaturated materials may dissipate faster than that in a larger
wetting zone within the unsaturated dam materials and so a
dam with low pore-water pressures or smaller wetting zone
is more stable than that with high pore-water pressures andlarger wetting area. As a result, the upstream slope ofsaturated dam materials is more stable than that ofunsaturated dam. However, it is noted that defining saturated
conditions for the dam materials may lead to inaccurate
results. In the real situation earth dam materials such as clay
core, embankment and sand filter are not fully saturated eventhough the reservoir level is level to the dam crest.
4.2 Effects of Drawdown Rate on the Slope Stability
To investigate the effects of drawdown rate (R) on thestability of the upstream slope during the reservoir
drawdown, the method of analysis described in section 3 wasrepeated for various drawdown rates (e.g., 8 m/day, 6 m/day
2 m/day and 0.5 m/day). The results of the analysis areshown in Fig. 7.
4.4

R=0.5 m/day
R=2 m/day
R=6 m/day
R=8 m/day

4

FOS

3.6
3.2

2

24

22

20

18

16

14

12

10

The effects of saturated hydraulic conductivity ofembankment soil and the drawdown rate of the reservoir on
the stability of the upstream slope of an earth dam subjected
to reservoir drawdown were investigated incorporating
unsaturated soil properties. The following conclusions weredrawn from this analysis:• Stability (FOS) of the upstream slope of an earth dam
decreases with the drawdown of the reservoir andreaches the minimum FOS at a reservoir water level ofabout 2/3 of the total drawdown height. Further decreasein the reservoir water level induces a gradual increase in
the FOS of the upstream slope.
• During the drawdown of reservoir, the upstream slope ofthe earth dam constructed with highly permeable
embankment soil is more stable than that of the dam withlow permeable embankment soil.

• A low rate of reservoir drawdown increases the stabilityof the upstream slope of the earth dam.
• Throughout the drawdown regime, the stability of the
upstream slope calculated using the saturated shear
strength of the dam materials is slightly higher than that
calculated using unsaturated shear strength.

The authors wish to acknowledge the Australian Agency forInternational Development (AusAID) for providing ascholarship for the first author’s studies at Queensland
University of Technology (QUT).

ksat = 3x10-7 m/s
26

5. CONCLUSION

6. ACKNOWLEDGMENT

2.8
2.4

consequently the slope stability increases. By contrast, thehigher drawdown rate or rapid drawdown (e.g., 8 m/day)will not allow enough time for pore-water pressure
dissipation. Therefore, the shear strength increase of the dam
materials subjected to a higher drawdown rate is less thanthat of the materials at a lower drawdown rate.

7. REFERENCES

8

Reservoir drawdown (m)Fig. 7. Variation of FOS of upstream dam slope withdifferent drawdown rates
As shown in Fig. 7, the FOS of the upstream slope of thedam increases as the drawdown rate decreases. The low
drawdown rate allows more time for the pore-water pressure
dissipation from the saturated zone within the dam.
According to Eq. (2), the dissipation of pore-water pressure
increases the shear strength of the dam materials. The porewater pressure within the materials can reduce to a negativevalue (suction) during the dissipation process, and[1] Berilgen MM, "Investigation of stability of slopes
under drawdown conditions," J. of Computers andGeotechnics, vol. 34, 2007, pp. 81-91.
[2] Desai CS, "Drawdown analysis of slopes by numericalmethods," J. of the Geotechnical Engineering Division,
ASCE, 103 (GT7), 1977, pp. 667-676.
[3] Fredlund DG and Xing A, "Equations for the soil-water
characteristic curve," J. of Canadian Geotechnical
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[4] Fredlund MD, Wilson GW and Fredlund DG, "Use of
the grain-size distribution for estimation of the soilwater characteristic curve," J. of Canadian
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[5] GEO-SLOPE International Ltd., Seepage Modelingwith SEEP/W 2007: An Engineering Methodology (4thed.). Alberta: GEO-SLOPE International Ltd., 2010.
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Slopes under Drawdown Conditions," J. of
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during rapid drawdown," Geotechnique, vol. 13, 1963,
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512

Lateral stress induced due root-water-uptake in unsaturated soils
Mu’azu Mohammed Abdullahi1 and Nazri Bin Ali2
Department of Civil Engineering and Built Environment
RAZAK School of Engineering and Advanced Technology,
Universiti Teknologi Malaysia, 54100 Kuala Lumpur Campus, Malaysia.
2
Department of Geotechnics and Transportation, Faculty of Civil Engineering, Universiti Tecknologi Malaysia
81310 UTM Skudai, Johor Bahru, Johor, Malaysia.
Email: 1muazu@ic.utm.my, 2nazriali@utm.my
1

ABSTRACT
A two-dimensional (2-D) lateral stress was modeled as a result of matric suction change
caused by vegetative induced moisture transfer. The negative pore-water pressures are
estimated through governing partial differential equations for unsaturated soils. The results of
the of simulated root water uptake are used as an input for the prediction of 2-D lateral stressin a stress-deformation analysis in an uncoupled manner. The soil is allow to expand andcontract free laterally, as the as water is being abstracted from the soil. A mature Lime treelocated on a Boulder clay sub-soil for period covering a full spring/summer drying period wasused as a case study. The result shows interdependence of lateral and vertical stress generated
resulting from root water-uptake.
Key words: Unsaturated soils, numerical, water-uptake, simulation, stress-deformation, experimental.
displacements as a result of vegetative induced
matric suction changes.

1. INTRODUCTION
When a material is compressed in one direction, it
usually tends to expand in the other two directions
perpendicular to the direction of compression, thisphenomenon is known as the Poisson effect. The
coefficient of lateral stress is a measure of themagnitude of the Poisson effect. In general soil
mechanics, the deformation of soil matrix isanalyzed as a result of increase in applied load,
whereas in groundwater field, the soil deformationis studied due to extraction of groundwater. In bothcases, a soil displacement takes place, because porevolume decreased [1]. Withdrawal of water byplant roots results in change in water pressures andmoisture content in the soil. Soil settlement occurs
whenever there is an increase in effective confining
stress. In prediction of soil movement two
fundamental stages are generally involved; anassessment of the changes in moisture conditions
and the knowledge of the volumetric strains
induced by these change. A horizontal and vertical
distribution of roots determines the dispersal ofroot water-uptake [1]-[2]This paper employed two-dimensional axisymmetrical finite element approach to solve thetransient partial coupled flow and stressdeformation equations. The study was based oncase study of mature single lime tree on a Boulder
Clay as reported by [3], with the following
objectives; to employ two-dimensional axisymmetrical finite element approach to solve thetransient partial coupled flow and deformation
equations and to simulate the water-uptake anddeformation. The capillary potential was estimate
as a result of the root water-uptake was partial
coupled to estimate the lateral and vertical

2.

THEORETICAL FORMULATIONS SOIL
MOISTURE MOVEMENT
The first step is dealt with through the use ofmodified Richard equation [4]; two-dimensional
axi-symmetric governing equation for unsaturated
soils with sinks term:

∂θ ∂ψ
∂ 
∂ψ  1 ∂ψ
+ K+⋅=  K (ψ )
∂ψ ∂t
∂z 
∂z  r ∂r∂ ∂ψ  ∂K (ψ )K (ψ )+− S (ψ , r , z )

∂z
∂r 
∂r 

(1)

Where K(ψ) is the unsaturated hydraulicconductivity, t is the time, r and z are thecoordinate, θ is the volumetric moisture content
and ψ is the capillary potential, S(r, z) is the root
water extraction function and r is the radial
coordinate.
The root water-uptake extraction function is thesink term S(ψ, z, r) in the Equation 1; for wateruptake in two-dimensional axi-symmetric form [5].
Comprising of vertical and radial components
incorporating water stress function when soil
moisture is limiting:S (ψ , z , r ) =

513


4T
z 
r
α (ψ ) 1 −  1 −  (2)z r rr
 z r   rr ( )Where α ψ(dimensionless) is a prescribedfunction of the capillary potential referred to aswater-stress function.

Mature Tree
400 elements
1281 nodes

The numerical solution of Equation 1 was achieved
via the finite element spatial discretization
procedure and a finite-difference time-stepping
scheme adopting particular Galerkin weighted
residual approach. The parabolic shape functions
and eight-node isoperimetric elements are
employed [6].

10 m

3

GROUND DISPLACEMENTTHEORETICAL FORMULATIONThe second step is tackled through stressdeformation formulation considering unsaturated
soil mechanics concept using ground water field
concept. Constitutive relationships are tocompliment governing flow equation, thus,
providing
additional
relationshipbetweendeformation and stress state variables. A change in
the negative pore-water pressure occurs as a resultof root water-uptake and can be related to changes
in soil volume through the use of constitutive
relations. Swelling in the field occurs along therebound curve at an overburden pressure.
Shrinkage occurs along either a recompression
curve or the virgin compression curve. While thesoil is a normally consolidated clay with aconsolidation behavior that can be described by; σ v + ∆σ v − u wf
de = C r ln

(
)
(
)
−
+
σ
u
u
u
−
v
a
a
w
e 


10 m

Figure 1 Finite element mesh (10 m x 10 m)
The simulation employs a time-step size of 21600
seconds, which was held constant for the entire
period considered. The lower boundary of thedomain and the far-field vertical and horizontal
boundaries remained unconstrained (natural)
throughout the simulation. The soil parameters areshown in Table 1 for Boulder clay. Based on thefield observations provided by [3], the root zone ofmatured Lime tree is assumed to extend to a depthof 2.0 m and a radial distance of 5.0(3)

Table 1 Parameters used in the analysis for Case I
Parameters Values
Referencesks
10-6 m/s[3]Ta
5 mm/day[3]
ψd1500 kPa[8]γ19.65 kN/m3 [9]
e0
1.25
[10]
Cr
0.023
[11]
µ
0.30
[12]
θr
0.1
[5]
θs
0.55
[5]
α
0.028
[5]
m
0.29
[5]
n
1.4
[5]
l
0.5
[5]

Where de is the change of void ratio in the element,
C r is the re-compression index, σ v is the vertical
total stress, Δσ v is the change in the total verticalstresses, u wf is the final pore water pressure, and(u a -u w ) e is the matric suction equivalent [7].
4DISCRETIZATION, BOUNDARY AND
INITIAL CONDITIONS
The mesh consists of 8-noded isoperimetric linear
strain quadrilateral elements. The entire finite
element mesh consists of 1281 nodes and 400
elements; the axi-symmetric domain is shown in
Figure 1. The mesh was configured to offer somerefinement within the root zone area, since this is
the region where the most significant moisture
content variations were expected to occur.. Spatial
discretisation has been achieved via the finite
element mesh shown in Figure 1.

Where K s is saturated hydraulic conductivity, T a
is actual transpiration rate, Ψ w is suction at wilting
point, γ is unit weight of soil, e 0 is initial void ratio,
C r is re-compression index, µ is Poisson’s ratio, θ r
is residual water content, θ s is saturated watercontent, α is water stress, m is empirical shapefitting parameters, n is empirical shape fitting
parameters and l is soil specific parameter
generally assumed to be 0.5. The required soil
moisture retention characteristics and unsaturated514

hydraulic conductivity was simulated from theclosed form equation developed by [13].
0.05CASE STUDY ON MATURE LIME TREEON BOULDER CLAY
The particular experimental data is based on thefield measurements undertaken at a site located at
Stacey Hall, Wolverton, England [14]. The caseconsidered here relates to a single mature Limetree, 15 m in height, located on a Boulder clay subsoil. A uniform initial value of capillary potentialof -17 cm was applied throughout the domain;
representing an initial volumetric water content of
37.5 % which corresponds to a degree of saturationof approximately 93.75 % was used as an initialvalue of capillary potential. The initial value ofcapillary potential would be in steady state and thesubsequent steady state value is applied to thesimulation. The drying phase was represented via
the application of the transpiration rate.

1.0

Depth (m)

2.0
3.0
4.0
5.0

0.0

7.0
Fig. 3 Variation of vertical ground displacementwith depth at various lateral distances from Limetree after 30 days.
The variations of lateral and vertical grounddisplacement with depth at various lateral distancesfrom Lime tree after 190 days are shown in Figs. 4
and 5. Fig. 4 show lateral ground displacement for
190 days simulation, a zero lateral displacement
was recorded at zero lateral distance from the Limetree. 18.67 mm, 19.34 mm and 6.76 mm lateralground displacements at 1.4 m, 3.0 m and 4.9 m
away from the lime tree respectively are simulated.
At distance of 7.5 m and 10.0m away from Limetree, -12.23 mm and -15.11 mm lateral grounddisplacements are simulated. This means at the
distance of 7.5 m and 10.0 m away from the lime
tree the effect of root water uptake is negligible.
This shows that there is likely expansion instead oflateral shrinkage of the Boulder Clay.

Lateral Ground Displacement (mm)
-1
0
1
2
3
4

Depth (m)

1.0
2.0
3.0
4.0
5.0

0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

6.0

6 Results and Discussions
The graphs of variation of lateral and verticalground displacement with depth at various lateral
distances from Lime tree after 30 days are shown in
Figs. 2 and 3. Lateral ground displacements aredenoted with positive signs while the verticalground displacements are denoted with negatives
signs.
-2

Vertical Ground Displacement (mm)-8 -7 -6 -5 -4 -3 -2 -1 0 1

0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

0.0

Lateral Ground Displacement (mm)-20 -15 -10 -5 0 5 10 15 20 25

1.0

6.0

Depth (m)

2.0

7.0

3.0

Fig. 2 Variation of lateral ground displacementwith depth at various lateral distances from Limetree after 30 days.

4.0
5.0

The vertical displacements seem to be lager inmagnitude than the lateral displacements. For 30days simulation 7.13 mm vertical grounddisplacement at zero lateral distance from the Limetree was achieved compared to zero lateral grounddisplacement at the same lateral away from thelime tree. The result shows that there is no lateralground displacement beneath the Lime tree.

0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

6.0
7.0
Fig. 4 Variation of lateral ground displacementwith depth at various lateral distances from Limetree after 190 days.

515

The vertical ground displacements are shown in
Fig. 5. Vertical ground displacements -79.16 mm, 48.70 mm, -29.64 mm and -6.58 mm at 0.0 m 1.4
m, 3.0 m and 4.9 m away from the lime tree
respectively are shown in Fig. 5. At distance of 7.5
m and 10.0m away from Lime tree, 6.70 mm and6.11 mm vertical ground displacements are
simulated. Similarly, its means at the distance of
7.5 m and 10.0 m away from the lime tree theeffect of root water uptake is negligible. Thisshows that there is likely heave instead of verticalground displacement of the Boulder Clay.

0.0

Vertical Ground Displacement (mm)-160 -130 -100 -70 -40 -10 200.0
1.0
Depth (m)

2.0
3.0
4.0
5.0

Vertical Ground Displacement (mm)-90 -70 -50 -30 -10
10

6.0
7.0
Fig. 7 Variation of vertical ground displacementwith depth at various lateral distances from Limetree after 270 days

1.0
Depth (m)

0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

2.0
0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

Lateral and vertical ground displacements withdepth at various lateral distances from Lime tree
after 270 days are shown in Figs. 6 and 7. The
sequence is the same with Figs. 2, 3, 4 and 5, bothvertical and lateral displacement increase with anincrease in elapse time and decreases with depth.

7.0
Fig. 5 Variation of vertical ground displacementwith depth at various lateral distances from Limetree after 190days.

7. CONCLUSIONS
The approach proposed utilizes radial symmetry
and a linear distribution of water extraction ratewith both depth and radius. The results of the root
water-uptake analysis are then used as an input forthe prediction of displacements in a stressdeformation analysis in an uncoupled manner. A
mature Lime tree located on Boulder clay sub-soilcovering a full spring/summer drying period wasconsidered for the case study. The majority of the
moisture extraction occurred near the surface;
likewise, the ground displacement occurred mostly
near the ground surface. Ground displacement
reduced significantly when the distance from thetree increased. Both lateral and verticaldisplacements are time and space dependent. The
accuracy of the lateral ground displacement
simulation depends on accurate determination of
the magnitude of the Poisson effect.

3.0
4.0
5.0
6.0

0.0

-25

Lateral Ground Displacement (mm)-15 -55152535

Depth (m)

1.0
2.0
3.0
4.0
5.0

0.0m
1.4m
3.0m
4.9m
7.5m
10.0m

6.0

8. REFERENCES[1] Ali N, Mu’azu MA, Kamarudin A (2010). 2-D
Axi-Symmetric of Root Water-Uptake. The 8thInternational Confereence on Geotechnical andTransportation Engineering, Sustainability inGeotechnical andTransportation
Engineering, (GEOTROPIKA 2010) 1-3
December 2010. Kota kinabalu, Malaysia, pp.
1-5.
[2] Mu’azu MA, Ali N, Ahmed K (2010).
Rainfall evaluation of root water-uptake

7.0
Fig. 6 Variation of lateral ground displacementwith depth at various lateral distances from Limetree after 270 days516

induced deformation, ARPN Journal of Eng.
and Applied Science, 5(8): 32-41.
[3] Biddle PG (1998). Tree Root Damage to
Buildings, Pattern of soil Drying in Proximity
to Trees on Clay Soils. 1st Edn., Vol. 2,
Willow mead Publishing Ltd, Wantage, pp:
88-93.
[4] Richards LA (1931). Capillary conductance ofliquids through porous mediums, Physics 1,
pp. 318-333.
[5] Rees SW, Ali N (2006). Seasonal water uptake
near trees: A numerical and experimental
study. Geomechanics and Geo-Engineering,
1(2): 129-138.
[6] Zienkiewicz OC, Taylor RL (1989). The Finite
Element Method. 4th Edn., Vol. 1, McGrawHill, USA., pp: 648.
[7] Fredlund DG, Rihardjo H (1993). Soil
Mechanics for Unsaturated Soil. 1st Edn., John
Wiley and Sons Inc., New York, pp: 346-373.
[8] Feddes RA, Kowalik PJ, Malink KK, Zaradny
H (1976). Simulation of field water uptake byplants using a soil water dependent root
extraction function. J. Hydro, 31: 13-26.
[9] Indraratna B, Fatahi B, Khabbaz H (2006).
Numerical analysis of matric suction effects oftree roots." Geotechnical Engineering,
159(2):77 – 90.
[10] Samuels SG (1975). Some properties of theGault Clay from Ely-Ouse Essex water tunnel,
Geotechnique, 25(2): 239-264.
[11] Ng CWW (1998). Observed performance ofmulti-propped excavation in stiff clay, J.
Geoenv., 124(9): 889-906.
[12] Almeida MSS, Britto AM, Perry RHG (1986).
Numerical modeling of a centrifugedembankment on soft clay, Canadian
Geotechnical Journal, 23(20): 103-114.
[13]vanGenuchten MT (1980). A closed-form
equation for predicting the hydraulicconductivity of unsaturated soils. Soil Sci. Soc.
Am. J., 44(5): 892-898.
[14] Bureau of Meteology (2006) United Kingdom
Bureau of Meteology Meteorological Rainfall
Data (1979-1980), Meteorological Office,Exeter, United Kingdom, 2006

517

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The effect of pre-tensioned rock bolts on seismic behavior of urban tunnels
Farshad Fayyaz Jahani 1, Mohammad Sharifipour 2
Department of Civil Eng., Univ. of Razi, Kermanshah, Iran

ABSTRACTThe distribution of seismic waves from the bedrock through the soil layers can cause severe damages not only to thestructures on the soil surface but also to the subsurface structures. Previous studies have indicated the significant role ofthe site condition in modifying and changing the characteristics of the strong ground motion data. The importance of the
pre-tensioned force of rock bolts has been recognized by many researchers. But the studies of seismic behavior of
pre-tensioned rock bolts have been considered by fewer researchers. To investigate the impact of seismic behavior of
pre-tensioned rock bolts on force redistribution and displacement around urban tunnels, a numerical analysis was
carried out using finite element method. The analysis performed by considering three actual ground motion records
representing seismic motions with low, intermediate and high frequency content. The results show that the
pre-tensioned rock bolts have a minor effect on decreasing the total displacement on the urban tunnels in all seismic
content and intermediate influence on the lateral force decrease.
Keywords: pre-tensioned rock bolts, seismic behavior, urban tunnels, finite element method
1. INTRODUCTION
When a roadway is introduced in a rock mass, its natural stateof stress is disturbed locally as the rock mass attains a newstate of equilibrium. These redistributed or induced stresses
have a definite zone of effect within which rock failure may
occur, creating a zone of disturbed or failed rock whichshould be supported [1]. In urban subway, rock bolts havebeen introduced in recent years and the mechanism of rockbolting has been studied by a great many scientists [2].
Consideration of earthquake ground motions, the way they
propagate through the earth, their characteristics description
at a certain location and methods for incorporating thisinformation into engineering designs have been the subject ofconsiderable research and interest so far. The energy released
from a source mechanism will travel in the form of seismic
waves through the rock formation where some energy
absorption takes place. The seismic waves will be reflected
and refracted as they approach the zone of special interest.
Numerous studies on different earthquakes where site
amplification caused substantial damage and collapse ofmany buildings are available (Romo and Seed 1986).
Observations made after the destructive earthquakes haveshown a correlation between damage and local geology, withthe destruction being larger on soft soil than hard soil or rock
(Duck 1958; Medvedev 1965; Seed, 1986) [3]. Dynamic
effects on underground structures have often been neglectedbased on the assumption that their response to earthquake
loading is relatively safe as compared to that of surface
structures. Nevertheless, several examples of recorded

damage to underground structures for which seismic forceswere not considered in the original design can be quoted. For
example, Hashash et al. [1] describe the collapse of the
Daikai subway station in Kobe during the 1995 HyogokenNambu earthquake, the damages to highway tunnels inCentral Taiwan during the 1999 Chi-Chi earthquake and thecollapse of the Bolu tunnel in Turkey during the 1999
Koceali earthquake. Another example is the south-Hyogo
earthquake of 17th Jan 1995 did serious damage to subway
stations in Kobe. So investigating the effects of earthquakes
on underground structures and their behavior during
earthquakes has been the subject of many studies so far(Konagai 1998, Hashash et al 2001). In this study the
analysis performed by considering three actual groundmotion records representing seismic motions with low,intermediate and high frequency content of a/v (pick groundacceleration in g to pick ground velocity in m/s) so as todiscuss the effect of frequency content on seismic responseof pre-stress rock bolt on displacement and force on theurban subway tunnel. The principal object of rock bolting isto help the rock mass to support itself. Bolts often act toincrease the stress and the frictional strength across joints,
encouraging loose blocks or thinly stratified beds to bang
together and act as a composite beam [4]. Rock bolts arealways installed with tension and the pre-tensioned effecthas been studied by a number of researchers and most results
are reported to be positive. Firth and Thomas argued that
active preloads modify roof behavior by dramaticallyreducing bed separation and delimitations in the immediate
0.5–0.8 m of the roof [5]. Peng stated that resin-assisted
point-anchor tensioned bolts can be used to clamp thinly
laminated roof beds into a thick beam that is more resistant
to bending [6]. Stankus and Peng added that by “increasing518

frictional resistance along bedding planes, roof sag anddefection is minimized and the lateral movement due tohorizontal stress is unlikely to occur” [7]. Kang and Zhang
also stated positive attribudes on the pre-tensioned rock bolts[4]. Generally, pre-tensioned rock bolts are said to be more
efficient than bolts without pre-tension, because a stronger
beam can be built with the same bolt by utilizing a larger
installed load. But in these all investigations researcher
consider only the static behavior of rock bolt and therefore
seismic behavior of pre stressed rock bolt is less or not taken
into consideration. Therefore in this study we tend to describe
the effect of pre-tensioned rock bolt in three different of
frequency content and three pre-tensioned force on theseismic behavior of urban tunnels.
2

GENERAL PROPERTIES OF THE SIMULATEDMODEL AND OTHERS

A circular tunnel is considered with 6 (m) interior diameterand the 30 (cm) thickness the center of model is located 16(m) below the ground surface, is selected as the reference
underground structure for the present case study. The lining
is assumed to be composed by 0.35m thick precast concretesegments characterized by the following linear visco-elastic
parameters as shown in table [2].
In the analyses, first, the primary state (static analyses) isdefined then the dynamic analysis is performed. soil and theconcrete body of the tunnel are both modelled using viscoelastic constitutive model with consideration of Rayleighdamping coefficients of the materials in general, when
linear-elastic or linear-visco-elastic models are employed todescribe the soil behavior, particular effort should be put inthe selection of the stiffness and damping parameterscompatible with the effective deformation level expectedduring the earthquake. The physical properties andmechanical parameters of both materials are reported inTable [1]. The assumed profile of the small-strain shear
stiffness G0 with depth (Fig [1]) was calculated adopting therelationship proposed by Viggiani and Atkinson used in thisstudy [8].

In the present study, all of the models are subjected to threedifferent natural ground motion records with different
frequency contents representing seismic motions with highintermediate and low magnitude of a/v (pick groundacceleration in g to pick ground velocity in m/s) classified as
Hav, Iav and Lav, respectively, as shown in figures [2],[3]and [4]. The earthquakes' properties are presented in Table [1]including their magnitude, maximum acceleration, velocity,
time duration, their classification, and location of theiroccurrence. In order to investigate the effect of pre-tensionedrock bolts on the seismic behavior around the roadway tunnel,
five similar models were constructed in our present study. All
model parameters, except the tunnel support patterns, weresimilar in the five models. The first model was unsupported
and the second model was supported with no pre-tensioned
force, other three models were supported by pre-tensionedrock bolts with pre-tensioned forces (20-60-100) kN/m,
respectively. On the other hand, the number of bolts used inthis study is 4 and 8 that have been used as a symmetric andspacing from each other in Z direction is selected 1 m.
Table [1].Earthquakes properties

Fig [2]. Selected accelerogram for hav earthquake

Fig [3]. Selected accelerogram for Iav earthquake

Fig [1]. Profile of small-strain shear stiffness G 0

519

Fig [4]. Selected accelerogram for Lav earthquake.
Table [2]. Material properties used in this analysis.

3 FINITE ELEMENT MODELINGThe length, width and height of the model were 150 m, 1 m
and 60 m respectively. A circular roadway (internal diameter
6 meter) was placed in the center of model is located 16 (m)below the ground. The coupled numerical analyses were
performed with the finite element code PLAXIS2D [9], a
two-dimensional (plane strain and axi-symmetric) FE codethat implements the coupled Biot dynamic equations [10],adopting the so-called u–p simplification (where u is theskeleton displacement and p the pore pressure), assuming as
negligible the fluid acceleration relative to the solid skeleton.
The code adopts the Generalised Newmark method [11] forthe time integration under dynamic conditions. In this study,
the following standard values of the Newmark’s constants
were selected in all the analyses illustrated in thispaper:α N =0.3025 and ß N =0.6. Those values ensure that thealgorithm is unconditionally stable, while being dissipative
only for the high- frequency modes. In the dynamic solution,
the code allows to introduce frequency-dependent viscousdamping by means of the Rayleigh formulation, the damping
matrix being defined as follows:[C]= α R [M]+ ß R [K]Where w n and w m are the angular frequencies related to thefrequency interval f n /f m in which the viscous damping isequal to or lower than D.
The boundary conditions adopted for the static stages of theanalyses were the standard ones: nodes at the bottom of themesh were fixed in both vertical and horizontal directions,while those along the lateral sides were only fixed in thehorizontal direction. In the dynamic analyses, the bottom of
the model was assumed to be rigid and the lateral sides werecharacterised by the viscous boundaries proposed by Lysmer
and Kuhlmeyer with parameters a=1 and b=0.25
Suitable constitutive elasto-plastic relations were chosen
to represent the mechanical behavior of the soil and concrete
body of the tunnel using the Mohr-Coulomb failure criterion
which has been adopted throughout the PLAXIS2D analysis.
In addition to its robust constitutive models, PLAXIS2D also
includes various structural support elements. The structuralelement called ‘node to node anchor’ represents rock andsoil support as an axial force along a line and this approachsuffices for most soil and cable bolts under practicalconditions. In this study the length bolts are selected 3
meter. The grout around the node to node anchor is modeledwith geogrid that can be tolerated only tension. Impervious
interface elements were also introduced to model theinteraction between the lining and the soil, soil and node tonode anchor according to the formulation summarised in the
manual of the code. In particular, the interface wascharacterised by values of the shear strength parameters
equal to 0.7 those of the surrounding soil. The mesh
employed in the present study is reported in Fig [5]: it ischaracterised by a width equal to eight times its height, inorder to minimise the influence of boundary conditions on
the computed results. The domain was discretised in a total
number of 1580, 15-node plane strain triangular elements. In
the central part of the mesh, where the tunnel is located, thecharacteristic dimension of the elements h always satisfies
the condition:h≤h max = V s /0.857f max
R

R

R

R

R

Where V s is the shear wave velocity and f max is the
maximum frequency of the seismic signal.
In this study in order to demonstrate displacements andlateral forces, two nodes A and B are selected located on thelateral and top of tunnel, respectively.
R

R

R

R

(1)

Where M and K are the mass and the stiffness matrix of thesystem, respectively. The coefficients α R and ß R are obtained
considering the following relationship with the damping ratio
D [12]:Fig [5] Mesh geometry in the FE analyses and location of nodes A and B

520

4 RESULTS AND DISCUSIONFirst, for example, we show the deformed mesh andchanging in the position of the tunnel for (Iav) earthquake.

Fig [6]. Final deformed mesh and change the position of the tunnel for Iavearthquakeon the tunnel. In intermediate frequency content earthquake
and with number of bolt 4, the pre-stress force don't effecton the displacement on prescribe nodes and have a minor
effect on the applied lateral force but with increasing in prestress force, it is reduced. But in high frequency content
earthquake the obtained results is differ from other two typeof earthquake. In other words the effective of pre-stress rockbolt is appearing in high frequency content earthquake. With
comparison between Fig [15] and Fig [21] can be said that inhigh frequency content earthquake, the applied lateral force
extremely decreased than intermediate frequency content
earthquake. Also in high frequency content earthquake due
to low period the behavior of model will be somewhat close
to static mode.

As shown in Fig [6], in the seismic analysis andoccurrences of earthquake due to more stiffness of tunnel tosoil, the deformation shape of the tunnel ground top surface
is shown as lump. In addition, by comparing figures 2 and 3
it can be said that the position of tunnel has changed and theshape of the tunnel, due to earthquake load, is an ellipse as
shown in Fig [6]. In order to consider the effect of pre-stressforce, three different rock bolt forces (20, 60 and 100 KN/m)have been modeled representing pre-tensioned. On the other
hand, influence of earthquake's frequency content has been
investigated on seismic behavior (displacement and lateralforce on specified points that located on right and top of thetunnel as shown in Fig [5]) of urban tunnels by considerationof forenamed ground motion records. Analyses are
performed in two separate parts: the free field analysis which
considers only the soil medium without the tunnel for
generating insitu stresses and soil-tunnel-rock bolt analysis
which includes the tunnel structure and rock bolt in the soil
medium. The results of these analyses are summarized into a
series of charts shown in Figures [7] – [24]. It should be
noted that in all figures shown in this study, we consider
only effect of pre-tensioned rock bolts on seismic behaviorof urban tunnels. This means that, after generating the initialstress, the total displacement sets to zero. All displacement
and lateral forces shown in all figures related to seismic
analysis.
The total displacement of node A and B in low frequency
content earthquake in no bolt mode is minimum and in withbolt mode (number of bolt 4), the total displacement of node
A is increased. On the other hand with increasing in prestress force the total displacement of node A and B
decreased. But the total displacement into with bolt mode ismore yet. With increasing in pre-stress force particularly in
100 KN/m, the applied lateral force on the node A decreases.
A little difference exists between two pre-stress forces 20,
60 KN/m.
In with bolt mode and number of bolt 8, in low frequency
content earthquake, the result is the same with the number ofbolt 4 but there is difference among different pre-stressforces on the total displacement in prescribe nodes. Note thatin this mode (number of bolt 8) increases in pre-stress force
have a minor effect on the reduction of occurred lateral force521

Fig [7]. Total displacement of node A in (lav) with number of bolt 4

Fig [8]. Total displacement of node B in (lav) with number of bolt 4

Fig [9]. Lateral force of node A in (lav) with number of bolt 4.

Fig [10]. Total displacement of node A in (lav) with number of bolt 8

Fig [11]. Total displacement of node B in (lav) with number of bolt 8

Fig [16]. Total displacement of node A in (Iav) with number of bolt 8

Fig [12].lateral force of node A in (lav) with number of bolt 8

Fig [17]. Total displacement of node B in (Iav) with number of bolt 8

Fig [13]. Total displacement of node A in ( Iav) with number of bolt 4

Fig [18].Lateral force of node A in ( Iav) with number of bolt 8

Fig [14]. Total displacement of node B in (Iav) with number of bolt 4

Fig [19]. Total displacement of node A in (Hav) with number of bolt 4

Fig [15]. Lateral force of node A in (Iav) with number of bolt 4

Fig [20]. Total displacement of node B in (Hav) with number of bolt 4522

constitutive model is used to consider the behavior of soil andthe concrete body of the tunnel with consideration ofRayleigh damping coefficients of these materials. According
to this assessment, the results show that the pre-tensionedrock bolts have a minor effect on decreasing the total
displacement on the urban tunnels in all low and intermediate(lav and iav) seismic content. But in high frequency content(hav), the result of total displacement between the different
forces pre-tensioned rock bolts model and with bolt model
differ from each other considerably. In other words in (hav)earthquake use of pre-tensioned rock bolt increases the total
displacement. But does not mean that the use of pre-tensioned
have a negative effect on tunnel displacement. The appliedlateral force in low and high frequency content earthquake isless than the intermediate earthquake. Therefore, the use of
pre-tensioned rock bolts is efficient in (lav) and (hav)earthquake for decreasing lateral force that applied on theurban tunnel. Also the use of model with the number of bolt 8than the model with number of bolt 4 is more efficient.

Fig [21].lateral force of node A in ( Hav) with number of bolt 4

2 REFERENCES

Fig [22]. Total displacement of node A in (Hav) with number of bolt 8

Fig [23]. Total displacement of node B in (Hav) with number of bolt 8

Fig [24].Lateral force of node A in ( Hav) with number of bolt 8

5 CONCLUSION
In this paper, effect of pre-tensioned rock bolts on seismic
behavior of urban tunnels on total displacement and lateralforces of the tunnel lining have been evaluated by using finite
element method with PLAXIS software. For this purpose, thetunnel is considered with 6 (m) diameter and the center ofwhich is located 16 (m) below the ground surface. Then theeffect of dynamic loading with various frequencies content
earthquakes (high, intermediate, low) on the variations ofdisplacement and forces in described nodes in the tunnel
lining have been assessed. In the modeling, the visco-elastic[1] Karakus M, Fowell R J, “Back analysis for tunnelling
induced ground movements and stress reredistribution, ”
J. of Tunnelling and Underground Space Technology,
vol .6, 2005, pp. 514–524.
[2] Kang H P, Jiang T M, Gao F Q, “Effect of pre-tensioned
stress to rock bolting, ” J. of China Coal Society, vol. 7,
2007, pp. 680–685.
[3] Saleh Asheghabadi M, Matinmanesh, H, “Finite
Element Seismic Analysis of Cylindrical Tunnel in
Sandy Soils with Consideration of Soil-TunnelInteraction, ” in Proc. 12nd Int Conf. on StructuralEngineering and Construction, 2011, pp. 3162-3169.
[4] Mark C, “Design of roof bolt system,” New
Technology for Coal Mine Roof Support, Pittsburgh:
U.S. Department of Health and Human Services, 2000.
[5] Frith R, Thomas R, “ The pre-tensioning pedigree, ”
J. of Australia's Mining Monthly SupplementUnderground Equipment and Technology, vol.7, 1998,
pp.68–72.
[6] Peng S S, “Roof bolting adds stability to weak strata, ”
Coal Age Magazine, vol. 11, 1998, pp. 32–38.
[7] Stankus J C, Peng S S, “A new concept for roof
support,”Coal Age Magazine, vol. 9, 1996, pp. 2–6.
[8] Viggiani GMB, Atkinson HJ, “ Stiffness of fine-grainedsoils at very small strains. J. of Geotechnique ,vol.2,
1995, pp. 249–65.
[9] Plaxis 2D. Reference Manual, version 8, 2003.
[10] Biot MA, “ General theory of three-dimensional
consolidation, ” J. of Applied Physics , vol.12, 1941, pp.
155–64.
[11] Katona MC, Zienkiewicz OC, “ A unified set of single
step algorithms. III. The beta-m method, ageneralizationof the Newmark scheme, ” J. of
Numerical Methods in Engineering, vol. 7, 1985,
pp.1345–59.
[12] Clough R, Penzien J, “Dynamics of structures”. New
York: McGraw H

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Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect Of Silt Content On The Anisotropic Behaviour Of Silt-Sand
Mixtures2

Navid Khayat 1, Abbas Ghalandarzadeh
Ph.D. Assistance Professor, Department of Civil Engineering, Islamic Azad University Ahvaz Branch, Ahvaz,
Iran. (Khayat@iauahvaz.ac.ir)
Ph.D,Associate Professor, School of Civil Engineering, University college of engineering of Tehran, Iran,(aghaland@ut.ac.ir)
P1F

P

P2F

ABSTRACTThe anisotropy of sand has not been fully studied in terms of the effects of different silt content. This study investigates
the effects on anisotropy in Tehran sand with various silt contents. The tests performed in this research include some
Hollow Cylinder Torsion Shear Tests conducted on 400kPa confining stresses. The samples had 0 ,15, 30, and 50( )percent silt contents, and the effect of the inclination angle on the major principal stress α was investigated with aparticular emphasis on silt content. The parameter "α" is considered as the key parameter that indicates the anisotropycharacteristics, and can vary from 0 to 90 . According to the results, an increase of α leads to more contractive
behaviours in the studied sands. In all samples, increasing the silt content up to 50% causes the strength to decrease.
Keywords: Sand; strength; Silt content; anisotropy;

Parameter α  (the inclination of major principal stress withrespect to the depositional (vertical) direction) is used todenote the rotation of principal stresses andparameter b′ = (σ 2 − σ 3 ) / (σ1 − σ 3 ) indicates intermediate

1 INTRODUCTION
In recent years, researchers have performed extensive andcomprehensive studies on clean sand, determined different
aspects of its behaviour in terms of its structure, and havemade several general conclusions. However, the behaviour ofclean sand containing silt is associated with several
complications that make it difficult to understand.
Extensive studies have focused on the effects of inherent andinduced anisotropy on the behaviour of clean sand[14]-[17]-[19] ]-[34].
Yamamuro and Lade (1998) extensively investigatedthe behaviours of sands containing small amounts of silt(7%)[30]. They concluded that when a small amount of silt
is added to the host sand, the obtained sand-silt mixture
exhibits the opposite behavior of that of pure sand. The sandunder low confining pressure (25 kPa) is completelyliquefied, and the strength of sand reaches zero. As theconfining stress increases, the sand with 7% silt has morestable behaviour than that of pure sand with no liquefaction
under high confining pressures. According to Yamamure
and Covrt (2001), the behaviours of silt-sand mixtures withhigh amounts of silt (40%) are different from those of puresand and those of silt-sand mixtures with small amounts offine particles [31]. In these mixtures, like the sands with 7%
silt, liquefaction occurs completely under 25 and 50 kPa
confining stresses; however, as the confining stressincreases, the behavior tends to be more stable. Suchbehaviours are similar to that of sand with 7% silt content
but are still different from those of pure sand [31].

stresses. Different principal stress directions and b′ values
can cause changes in soil behaviour. Thus, α  and b′ areselected as the key parameters for studying the anisotropy ofsands. In this regard, Hollow Cylindrical Torsional Shear(HCTS) is the best apparatus for studying these parameters[1]. HCTS has been used in several tests conducted on sandsby different researchers to study the effects of parameters α and b′ more accurately. They generally came to theconclusion that sandy samples had softer behaviour as thevalues of α  and b′ increased in undrained tests [34].
According to Symes et al. (1985), sandy samples exhibit
more contractive behaviour as the α  and b′ values
increased[14]-[25].
Despite extensive studies conducted on the anisotropy effects
of clean sand, there are few studies that have investigated its
effects on the behaviour of sand-silt mixtures. Among recent
relevant studies, Bahadori and Ghalandarzade (2008) studied
the inherent anisotropic effects on the undrained behaviour ofsands using HCTS[4].
In this research, the anisotropic behaviour of Tehran sandwith different silt contents in 400 kPa initial effective
confining stress are studied. A total of 12 tests are conductedon Tehran sand with silt contents of 0%, 15%, 30% and 50%.
In these tests, the effects of the inclination angle of majorstress and silt content are investigated.

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2 Material and Methods

to 4 degree per minute. Eleven transducers were adjusted inorder to measure different parameters continuously .Threewater pressure sensors, two for inner and outer cell pressuresand one for pore pressure, one vertical displacement sensor tomeasure axial strain, and one rotation angel sensor tomeasure torsional strain, two differential pressure sensors,
two axial and torsional stress sensors and finally two limit
switch sensors were set up. A computer program was utilized
to control all test procedures by means of a PID control
algorithm. Data logging device was the last part of thecontrol system. The sample dimensions were 12 cm in height,
6 cm in inner diameter and10 cm in outer diameter.
The maximum inclination of principal stress ( α  ) andintermediate stress ratio ( b′ ) are considered the criticalparameters in this research. Because the torsion sensor
measures the torque (T) at any moment, the σ Zθ (shear

2.1 Material properties
Tehran sand, which is produced by crushing plants, This sandwith greenish gray color. The physical characteristics andgrading curves of this sands are shown in Fig. 1 and Table 1.
The plasticity index of silt is less than 5% and can beaddressed as non-plastic soil (see Table 2).
2.2 Maximum and minimum void ratios of sand-silt
mixtures
An effective factor in the general behaviour of saturated
sands in monotonic loading is soil relative density, which is aparameter that shows the sample’s specifications at the end of
consolidation stage. The samples prepared at low densities
showed contractive behaviours in the drained tests; thisbehaviour causes excess pore water pressure in undrained
tests, which reduces the effective confined stress. The changeof stress-strain from the softening to hardening states isgenerally related mostly to the relative density, which isobserved
andreportedby several researchers[5]-[7]-[8]-[16].
It is clear that adding silt to the host sand changes the
maximum and minimum void ratios. The parameters related
to soil density are needed to compare the test results;
therefore, minimum and maximum void ratios should beobtained for different percentages of silt. In this research,
ASTM 4253 and ASTM 4254 are used to measure theminimum void ratio ( emin ) (densest state) and maximum

stress) value can be obtained using Eq (3), where ri and ro
are the inner and outer radiuses of the samples, respectively.
The value of σ Z (vertical normal stress) must be kept
constant,

andtherefore,

thevalues

of

σθ

and

σ r (circumferential and radial normal stress respectively)
obtained through Eqs (1) and (2), respectively, are also kept
constant. Furthermore, Pi and Po (inner and outer waterpressures) are computed using Eq (5), Fν (vertical force) isfound using Eq (4) and principle stress can be measured byequation 6 to 8 [15].

void ratio ( emax ) (loosest state), respectively. The obtained
values are plotted in Fig. 2. For host sand, the minimum void
ratios decrease as the silt content of the samples increases.
This trend continues up 28 percentage fine content (FC);then, the void ratio increases as the silt content increases. The
minimum achieved void ratio takes place in the fine contents,
which will be called the threshold fines content.

σθ = σ Z −

(1)

2σ Zθ
tan 2ασr = σZ −1

σ Zθ (cos 2α − 2b + 1)
sin 2α
3T

T

(2)


(3)

+
σ Zθ = 2  2π (ro3 − ri3 ) π (ro3 + ri3 ) (ro − ri ) σZ =2.3 Testing apparatus and method
In order to study the inherent anisotropy, a hollow cylinder
torsional shear (HCTS) device was used. For better
investigation of post peak response the device can perform
both stress and strain controls.
A closed loop control was applied to the machine to handle
any type of desired stress history including major principal
stress rotation and intermediate stress ratio. FourElectrical/Pneumatic [E/P] transducers were used to exert
required pressure for the inner and outer cell pressure in
addition to the torsion and axial load pneumatic actuators. An
alternative DC motor for torsional strain control test was
prepared in order to study the post peak behaviour. In thecase of our tests, the motor was utilized since the strain
control was adopted in order to study the softening response.
The motor speed in all the tests remained constant and equal

FV + π ( Po ro2 − Pi ri 2 ) − Ar PoAS

σ r (ro + ri ) + σ θ (ro − ri ) Pi =
2ri


σ
(
r
r
)
+
 P = r o i + σ θ (ro − ri )
 o
2ro

(4)
(5)

2.4 Experimental Program
Sample preparation method used in this study was dry
deposition method. That may alter the real anisotropy fabric;
water sedimentation method was not used. Dry deposition
method is able to create more uniform silty sand samples
according to Miura and Toki (1982) suggestion which says
pluviation methods create the most significant anisotropy[12].The sand was deposited into the frame through a funnel
with a long tube. Carbon dioxide ( CO2 ) and deaired water
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4.1 Effect of inclination angle ( α ) on the sand’s
behaviours

are passed through the samples in their saturated states. The
circulating time for the CO2 considered in the tests
conducted by Zlatovic and Ishihara (1997) were 30 minutes
for clean sand and 8 hours for pure silt [35]. In the present
study, the minimum and maximum times were 30 and 135
minutes for clean sand and the 50% silt mixture,
respectively. After circulation of CO2 , the deaired water
entered from the bottom similar to CO2 but in the opposite
direction of gravity and was seeped into all the voids in thesample. The saturation procedure continues as the confiningpressure increased, and the pore water pressure wasmeasured in several steps. If the B-value would exceed 0.95,
then the sample is assumed to be completely saturated. At
this time the saturation stage was completed and the sample
should be consolidated. Because all the tests conducted inthis research were of the Consolidated Undrained (CU) type,all drainage valves must be opened during consolidation andconnected to the system. The amount of water discharged
from the sample during consolidation could be measured byreading the burette numbers at the beginning and end of thisstage. After consolidation the shear stage started at which
the shear speed was four degrees per minute, the lowest
speed that can be applied to the system. At the end of the test
the porosity of the samples was measured.



In Figs. 3 through 6, the strength of samples (deviator
stressing phase transformation and failure strength) is
reduced as α  increase, which indicates that the soil
becomes softer. This behaviour has been described by
Yoshimine and Ishihara (1998) for Toyoura sand and by
Bahadori and Ghalandarzadeh (2008) for a Firoozkuh sandsilt mixture[4]-[34]. It seems regardless of type of sand,similar behaviour can be observed in all studied cases here.
4.2 Effect of non-plastic fines content on the sand’s mixture
behaviours

3. Results
The goal of this research was to study the influence of theamount of silt upon anisotropy features in sand-silt mixtures.
For this purpose, sand were tested using 400-kPa initialconfining stress and a silt content range of 0-50%.Also theIntermediate stresses parameter (b′) in all tests was 0.5,
Some information on these tests are given in Table 3.
All test results are presented in the form of stress-strain andeffective stress path curves. According to these curves, thesample behaviours are similar to those of dense sands based on
the classification of undrained sand behaviour in mostconducted tests, according to Yoshimine et al. (1998) [34].
The significance of the strength of the deviator stressing phasetransformation point is evident in this figure. According to
Figs. 3 through 6, the sample strengths (deviator stressing
phase transformation and failure points) decrease as α increase, which indicate that the soil becomes softer.
Increasing the silt content up to the threshold level will cause
the strengths to decrease, and this reduction is more obvious
for the 50% silt case (Figs. 6). In these samples, the test
conditions were the same, and the silt percentage increased as
the density increased, but unexpectedly, strength did notincrease. For example, in test No.7, the value of Dr is 76, but
in test No.28, which contains 50% silt, its value is 99, 6.
4. Discussing and studying the stress paths

526

The effect of nonplastic fines on the behaviour of saturated
sand has been studied by many researchers [6]-[13]-[23][29]-[32]. Adding different amounts of nonplastic silt
considerably changes the sand behaviour, and these changes
are difficult to describe. Pitman et al. (1994) concluded thatwhen silt was added to Ottawa sand, it became lesscollapsible in undrained triaxial compression tests.
However, others found that nonplastic silt may decrease thestrength (Troncoso and Verdugo, 1985; Sladen et al., 1985)[18]- [24]. Ishihara (1993) and Verdugo and Ishihara (1996)
showed that the increased silt content increased the potential
for sand to exist in nature in a contractive state, which
includes the possibility of flow failure or liquefaction[9][28]. Yamamuro and Lade (1997, 1998) and Lade andYamamuro (1997) observed that increasing the amount of
nonplastic silt in Nevada sand increased the volumetriccontractive character in both drained and undrained triaxial
tests, even when the overall density increased[30]. Amongthe problems involved in these studies is the impossibilityof fixing at least one of the parameters in sand-silt mixtures.
In the present study, the method proposed by Bahadori and
Ghalandarzadeh (2008) [4] was applied to address thisproblem. Based on this method, if the same conditions are
considered in sampling and testing, then increasing the silt
increases the density and decreases the strength. Thus, while
the sample densities are equal, the sample with higher silt
content has lower strength.
The sample strengths decreased in the silt-sand mixtures as
the silt content increased (Fig. 7). The decreasing strength assilt is added to the host sand can be explained by
Tavanayagan’s classification (1999) (Fig. 8) [21]. According
to Fig. 9, the void ratios of host sands are tested in the rangeof Tavanayagam’s classification case (i); however, with 15%
silt content, the intergranular void ratio (ec ) approachedemax, HS (Fig. 9), which is more similar to that observed in
Tavanayagam’s classification case (ii). In this latter case, thefiner grains may support a coarser grain skeleton that isotherwise unstable. These grains act as a load transfer vehicle
between some of the coarse grain particles in the soil matrix,
while the remaining fine particles fill in the voids. In this

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur,
Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051
case, the shear strength is derived from a combinationof
D50 : Medium grain size
friction along coarse grain contacts and fine particles. In thisd : Diameter of fine grains

structure, ec increased, which decreased the strength of thesamples. This behaviour is evident as the silt percentage inthe mixtures increases and leads to an increase in relative
density. Increasing the amount of silt content to 30% caused
the intergrain void ratio increased, the relative density does
not increase because its threshold percentage was 28%. Thus,
an inter-fine-particle void ratio is used to explain its
behaviour, which is similar to case (iv-2) and shown in Fig.
9. Moreover, the inter-fine-particle void ratio in the mixture
with 30% silt content is greater than that of a sample with
15% fine particle content. When the silt content is increased
to 50% in all mixtures, e f is smaller than ( emax, HF ) ,and

ec : intergranular void ratioe f : Inter-fine-particle void ratio
emax , emin : Maximum and minimum void ratios, respectively
emax,HF , emin,HF : Maximum and minimum void ratios of hostfine particles, respectively
FC : Fine particle content (%)
FCth , FC L : Threshold and limiting fine particle content,
respectively(e1 − e3 ) / 2 : deviatoric axial strain( FCth < FC ); thus, their behaviours are similar to
Tavanayagan’s classification case (iv-2). In this case, the fineparticles carry the contact and shear forces, while the coarse
grains act as reinforcing elements embedded within the finer
grain matrix. In this situation, the main bearing structurechanges from sand to silt, and because the strength of silt is
less than that of sand, increasing the silt up to 50% decreases
the strength of the mixture.

pc′ : initial effective confining stresspi , po : inner and outer water pressures, respectively
ri , ro : inner and outer radius, respectively
6. REFERENCES

5. Conclusions
1) As α  increases, the strengths of the samples of host sandand silt-sand mixtures decrease, i.e., the soil becomes softer.
Such behaviours have also been observed by Yoshimine etal. (1998) and Bahadori and Ghalandarzadeh (2008) [34]-[4].
2) As the silt content in mixtures increases, the relative
density also increases, but this increase does not increase themixture strength. An increase in relative density is one of themost marked phenomena that indicate that the index of therelative density (or void) is not a suitable index forevaluating the behaviour of silt-sand mixtures.
3) In all mixtures, increasing the silt content up to 50%
causes the strength to decrease. This decline in strength may
be justified by parameters ec , e f , which are defined by()

Tavanayagam (1999) [21].
Notation
The following symbols are used in this paper:
α : the inclination of the major principal tress with respect tothe vertical directionb′ : Intermediate stressesb : Portion of fine grains that contribute to active intergrain
contacts
D : Diameter of coarse grains
Cc : Coefficient of curvature

Cu : Coefficient of uniformity
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Azad University,Tehran, Iran.
[11] Lade, P.V. and Yamamuro, J.A. (1997). Effects of nonplastic fines on static liquefacltion of sands. Candian
Geotechnical Journal, 34.No. 6, 918-928.
[12] Miura. S. and Toki, S. A.(1982). Sample preparation
method and its effect on static and cyclic
deformation-strength properties of sand, Soil andFoundation 22, No. 1, 61-77.
[13] Naeini, S.A. & Baziar, M.H. (2004). Effect of fines
content on steady-state strength of mixed and layered
samples of a sand.. Soil Dynamics and Earthquake
Engineering 24, 181-187.
[14] Nakata, Y., Hyodo, M., Murata, H. and Yasufuku, N.
(1998). Flow deformation of sands subjected to principal
stress rotation. Soils and Foundations 38, No. 2,
115-128.
[15] Pradhan, B. S., Tatsuoka, F. and Horii, N. (1988).
Simple shear testing on sand in a torsional shear
apparatus, Soils and Foundations 28, No. 2, 95-112.
[16] Seed, H.B. and Idriss, I.M.(1971). Simplified procedure
for evaluating soil liquefaction potential. GeotechnicalEngineering Journal 97, No.9, 1249-1274.
[17] Sivathayalan, S. & Vaid, Y.P. (2002). "Influence ofgeneralized initial state and principal stress rotation on
the undrained response of sands", Canadian
Geotechnical Journal;39:63-76.
[18] Sladenn J. A., D'Hollander, R. D. and Krahn, J.
(1985a).Back analysis of the Nerlerk berm liquefaction
slides. Canadian Geotechnical Journal 22, No.4, 564588.
[19] Symes, M.J., Gens, A. and Hight, D.W.(1984).
Undrained Anisotropy and Principal stress Rotation in
Saturated Sand. Geotechniqu 34, No. 1, 11-27.
[20] Symes, M.J., Gens, A. and Hight, D.W.(1985).
Discussion. Geotechnique, 35, No. 1, 78-85.
[21] Thevanayagam, S. (1999). Effect of fines and cofining
stress on undrained shear strength of silty sands. Journalof Geotechnical and Geoenvironmental engineering 124,
No.6, 479-491.
[22] Thevanayagam, S., Shenthan, T., Mohan, S. and Liang,
J. (2002). Undrained fragility of sands, silty sands and
silt.
ASCE,
Journalof
GeotechnicalandGeoenvironmental engineering 128, No.10, 849-859.
[23] Thevanayagam, S. & Martin, G.R. (2002). Liqueffaction
in silty soils-screening and remediation issues. Soil
Dynamics and Earthquake Engineering ,No.22,
1035-1042.
[24] Troncoso, J.H. & Verdugo .R.(1985).Silt content anddynamic behaviour of tailing sand .Proceeding of 11th
JCSMFE,
Balkema,Rotterdam,The
Netherlands,pp.1311-1314.
[25] Uthayakumar, M. & Vaid, Y. P. (1998). Staticliquefaction of sands under multi axial loading.
Canadian Geotechnical Journal 32, No.2, 273-283.
528

Table 1: Physical characteristics of sands
Sand type

Cc

Cu

FC (% )

D50(mm )Tehran sand

0.59

2.55

0.03

0.106
Fig. 2. Minimum and maximum void ratios for the Tehran
sand-silt mixtures

Table 2: Physical properties of silt
Plastic limit

Liquid limit

Plasticity index

(%)

(%)

(%)

22

24

2

Table 3. Characteristics of the tests
Test

Sand

FC

No.

type

(%)

7

T

8

α

e

Dr (%)

0

15

0.711

T

0

30

9

T

0

16

T

17

(e ) (e )
f

c

76.0

1.83

0.71

0.712

75.8

1.83

0.71

60

0.718

74.3

1.85

0.72

15

15

0.600

83.0

1.25

0.84

T

15

30

0.610

80.7

1.27

0.85

18

T

15

60

0.606

81.6

1.26

0.84

25

T

30

15

0.560

89.3

0.98

0.92

26

T

30

30

0.563

88.6

0.98

0.93

27

T

30

60

0.567

87.7

0.99

0.93

28

T

50

15

0.530

99.6

0.76

0.88

29

T

50

30

0.534

98.8

0.77

0.88

30

T

50

60

0.537

98.2

0.77

0.89

Fig. 3. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa (FC=0%).

Fig. 4. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa (FC=15%).

percent by weight finer (%)

100
90
80
70
60
50
40
30
20
10
0
0.01000.10001.000010.0000

Grain Size (mm)Fig. 1. Grain size distribution curves of Tehran sandFig. 5. Stress path and stress-strain curves in Tehran sandsubjected to an initial effective stress of 400 kPa (FC=30%).
529

Tehran Sand
2.5

2.0

Void Ratio

e-max
e-min

1.5

0%
15%
1.0

30%
50%

0.5

0.0
0

20

40

60

80

100

Silt Content (%)

Fig. 6. Stress path and stress-strain curves in Tehran sandFig. 9. Intergranular matrix phase diagram for Tehran Sand

subjected to an initial effective stress of 400 kPa (FC=50%).

Fig. 7. Stress path and stress-strain curves in Tehran sandswith different silt contents at α = 15.

Fig. 8.Intergranular and Interfine Matrix Phase Diagram(Thevanayagam,1999) [18]530Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Production and Properties of High Strength Concrete for Heightening
Concrete Dam in Sudan
1

Osama M. A. Daoud 1 and, H. S. Sagady2
Building and Road Research Institute, University of Khartoum, Sudan
2
Senior Material Engineer, SMEC International Co. Ltd., Sudan

ABSTRACTThis paper presents a part of an ongoing experimental laboratory investigation being carried out for production andcharacterization of high strength concrete (HSC) for heightening of an existing concrete dam in the middle of Sudan.
Brief description of the main features of the dam and concrete works is presented. Hundreds of trial mixes were
performed and tested using local Sudanese aggregates with addition of mineral admixtures (Silica Fume and Fly Ash)and Super plasticizers. Six grades of HSC (50, 60, 70, 80, 90, 100 MPa) had been success fully produced and theirmechanical properties were measured and documented. Statistical analysis of tests results was performed and simple
correlations were developed relating compressive strength to flexural and Splitting Strengths.
The results have offered an important insight for optimizing the rheological characteristics of HSC and permitted todevelop guidelines for optimum mix design methods for HSC from locally available aggregates in Sudan. The effect of
w/c ratio on strength of HSC was also highlighted. It is concluded that local concrete materials, in combination withmineral admixtures can be utilized in producing High Strength Concrete in Sudan.
Keywords: High Strength Concrete, Silica Fume, Fly Ash, Dams, Sudan.
1. INTRODUCTION
Concrete, a composite material consisting of aggregates
enclosed in a matrix of cement paste including possible
pozzolans, has two major components, cement paste andaggregates. The strength of concrete depends upon thestrength of these components, their deformation properties,
and the adhesion between the paste and aggregate surface [1].
In recent years, the construction industry has shown
significant interest in the use of high strength concrete(HSC), in applications such as dams, bridges and high risebuildings. This is due to significant structural, economic andarchitectural advantages that HSC can provide compared toconventional, normal strength concrete (NSC).
Although high strength concrete is often considered a
relatively new material, its developed has been gradual overmany years. Definition of the minimum strength value forhigh – strength concrete varies with time and geographicallocation depending on the availability of raw material and thetechnical Know-how, and the demand form the industry [2].
Starting form a value of 34 MPa in the 1950s in the UnitedStates moving to upper values, the ACI 363, 1999 [4] on highstrength concrete defines a value of 51 MPa (cube strength)
as a minimum value for high for high strength concrete.
Production of HSC may or may not require special materials,
but it definitely requires materials of highest quality and theiroptimum properties [3]. The production of HSC thatconsistently meets the requirements for workability andstrength development places more stringent requirements onmaterial selection than that for lower strength concrete (ACI

363R, 1999) [4]. However, many trial batches are often
required to generate the data that enables the researchers andprofessionals to identify the optimum mix proportions for
HSC [3].
2. GENERAL DESCRIPTION OF ROSEIRES DAM
Roseires Dam is located in Damazin city in the Blue Nile
State, about 520Km south-east of Khartoum.The Dam wasoriginally designed to be constructed in two phases, the firstphase was completed in 1966 to maximum water level ofAlexandria Datum (AD) 483 m with allowance for it to beraised to AD 493 m at a future date.
The Dam is a composite dam with a central 1000 m longconcrete buttress dam section that contains the deep levelsluices, gated spillways, power intakes for a service power
station and head works for future irrigation canals on eitherside of the river. The maximum height of the existing
concrete section is 68 m. It embodies specialized section fordeep sluices, spillway, west and east bank irrigation canalhead works, main power station (280 MW) and service
power station (2 MW). The Embankment Dam is 8500 mlong on the left (west) bank and 4000 m long on the right(east) bank. The maximum height of Embankment Dam is 30
m and the maximum width 230 m. Once raised, Roseires
Dam will have a combined length of 25 Km (Fig. 1). Table 1
shows the main features and important numbers of thedifferent components of the dam before and after theheightening. As one of the major development projects in
Sudan, this project was given a top priority in the strategicplanning of the Sudanese government in the last ten years.

531

The Accepted Contract Amount for Roseires Dam
Heightening was around three hundred and ninety-six million
United States Dollars (USD 396,000).
Time required for Completion of the Works is 1308 days,
making October 2012 as the completion date for the whole of
the works. All concrete works are expected to finish by theend of May 2012, while the rest of other electromechanicaland earth embankments works are expected to finish by
October 2012.
1.2 Main Concrete Works

Fig. (1): General Layout of Dam site
Table 1: Main Features of the Dam

Description
Existing Dam
After Heightening
Reservoir
Volume (without siltation) 3,000 million m3 7,400 million m3
Area
290 Km2
627 Km2
Maximum Water Level
AD 483.02m
AD 493.02m(FSL)
Overtopping Level (Concre
AD 485.22m
AD 495.22m
Dam)
Overtopping Level
AD 485.52m
AD 495.52m(Embankment)
Minimum Draw down
AD 438.52m
AD 438.52m
level
Tailwater
Maximum Water Level
AD 459.02m AD 459.02m (PMF
(PMF)
Nominal Wet Season
AD 451.02m
AD 451.02m
Tailwater level
Minimum Tailwater level
AD 443.02m
AD 443.02m
Contract Dam
Roadway levelAD 484.22m
AD 494.22m
Maximum level above
68.0m
78.0m
Foundation
Crest Length
1,000m
1,000mConcrete Volume
850,000 m3
155,000 m3
Embankment Dam
Crest Length (east)
4,000 m
8,500 mCrest Length (west)
8,500m
15,500m
Fill Volume
5 million m3
17 million m3

The major concrete works in the project included theheightening of the existing concrete dam by ten meters above
the existing concrete level (483.02 m AD) which will
accordingly increases the crest levels of both eastern andwestern embankments from 4000 m to 8,500 m and from8,500 to 15,500 m respectively. The heightening of theconcrete dam will increase the volume of the reservoir from
3,000 million m3 to 7,400 million m3 of water per year. A
total of 155,000 cubic meters of high strength concrete were
properly prepared, mixed, casted and cured using locallySudanese aggregates, flay ash and locally produced ordinary
Portland cement.
Heightening of the existing concrete dam required extension,
web thickening and strengthening of buttress andenlargement of the deep sluices (Fig. 2). In order to pour the
new high strength concrete on the existing old concrete,anchor bars and high tensile strength epoxy resin was used asa grout after demolishing and roughening the surfaces ofexisting concrete. Table 2 below shows the details of anchor
bars. Fig. 3 shows a general view of the dam when most of
concrete works were finished.

SMEC International in association with Coyne et Bellier ofFrance undertook a review of the tender design and tender
documents which were prepared during the early 1990,s andhas updated the design as necessary, including preparation of
new tender documents.
In January 2009, Lahmeyer International (LI) of Germany
substituted Coyne et Bellier as sub-consultant to SMEC. The
main contractor of the project was CCMD Joint Venture of
SINOHYDRO and CWE (two Chinese companies).

Fig. 2: Thickening and Strengthening of buttresses532

Fig. 3: General View of Concrete Dam Almost Finished
Aggregates:

Table 2: Details of anchor bars
AnchorBars
Diameter
(mm)
16
32

AnchorBars
Length
(mm)
1400
2800

Diameterof Drilledhole (mm)Length ofDrilledhole (mm)

25
45

700
1400

Spacing at
site
staggered
(mm)
1000
1500

3. EXPERIMANTAL PROGRAM
The following subsections present the details of the materials
used in the production of HSC and the related testing andspecifications.
3.1Concrete Ingredients
CementIn this research, a locally produced ordinary Portland cement
type I, conforming to ASTM C150 (OPC 42.5N) [5] which isextensively used in Sudan, was used in the trial batches
production. The specific gravity of cement used was 3.15,
initial and final setting time were 3:17 and 4:38, otherphysical and mechanical properties test for cement are shownin Table 3.

The coarse and fine aggregates used in this study werecrushed marble processed from the local quarries around
Damazin City, the quarry for Roseires Dam Heightening
Project. The maximum aggregate size was 20 mm, thegrading of the coarse and fine aggregates is shown in Figure
4. The specific gravity and absorption of the coarse
aggregates, determined in according with ASTM C127 [8]were 2.84 and 0.25respectively, whereas those of fineaggregates, determined in accordance with ASTM C128 [9]were 2.839 and 0.45 respectively. All the sand samples were
tested for their absorption percentage in saturated surface dry(SSD) condition. Organic impurities in sand were tested inaccordance with ASTM C-40 [10]. The water-cement rationof all trial mixes were based on saturated surface dry
condition (SSD) of the aggregates. For the mix HSC6,
different type of aggregates from another quarry was used.
To compare with marble, granite aggregates from Merwei
Dam (recently constructed another concrete dam in the northof Sudan) location were used.

Table 3: Physical and Mechanical Properties of Cement
Test
ResultNormal Consistency
29%
Setting Time
Initial

3:17 hrs

Setting Time
Final Setting

4:38 hrs
Fig. 4: Gradation of Fine & Coarse Aggregates

Time
Loss on ignition

2.29%

Compressive Strength
2 Days20.8 Mpa

28 Days56.4 Mpa

Chemical Admixtures (Superplasticizer)
The superplasticizer used in this study has the trade nameof “PCA-(I)” from Jiangsu Bote New Materials
Company-China.
PCA-(I)isapolycarboxylate
polymer-based composite admixture. It is a liquid which has
the performance of high range water reduction, excellent
slump retention and strengthening. The specific gravity of the533

superplasticizer was 1.085 and the PH was 8.11 with nil
chloride content percentage by weight. It is specially adapted
for the production of high durability concrete,self-compacting concrete, high compressive strength
concrete, and high workability concrete. PCA-(I)superplasticizer is formulated to comply with the ASTM
specifications for concrete admixture: ASTM494, Type G[11].
Mineral Admixtures (Silica Fume and Fly ash)
Silica Fume
Silica fume (SF) is a by-product of the manufacture ofsilicon metal and ferro-silicon alloys. The process involves
the reduction of high purity quartz (SiO2) in electric arc
furnaces at temperatures in excess of 2000̊̊C. SF is a very fine
powder consisting mainly of spherical particles or
microspheres of mean diameter about 0.15 microns, with avery high specific surface area (15,000-25,000 m²/Kg). Eachmicrosphere is on average 100 times smaller than an average
cement grain. At a typical dosage of 10% by mass of cement,
there will be 50,000-100,000 SF particles per cement grain.
In bulk, SF is generally dark grey to black or off-white incolour and can be supplied as a densified powder or slurry
depending on the application and the available handlingfacilities. SF is available globally. There is no health hazard
associated with the use of SF as non-crystalline silica is
non-hazardous. The dense microstructure of concrete
containing SF leads to major improvements in mechanicalperformance and resistance to chemicals (such as acids, fueloil, chlorides and sulfates). SF is ideally suited to the mostdemanding applications, such as concrete slipways, dam
spillways and hard standings, where chloride, chemical orabrasion resistance are required. SF concretes have
performed well under these circumstances, as they arechemically stable and have very low permeability. The SF
used in this study was in accordance with the mostinternational standards such the European BS EN 13263
Silica fume for concrete, Part 1:2005 Definitions,requirements and conformity criteria Part 2:2005 Conformity
evaluation, and the American ASTM C1240-97b Standard
specification for silica fume for use as a mineral admixture in
hydraulic- cement concrete, mortar and grout. The specificgravity of the silica fume silica fume used in this study was
2.373. SF the pozzolanic high activity, which can be filled
the gap between cement, increase the density of the system,so as enhance strength, impermeability, wear proof,
anti-corrosion, anti-scour, antifreeze, and strong early
performance.
Table 4: Physical Properties of Silica Fume
Moisture
Content%

Loss onignition
%

0.22

0.48

Fineness
Retain
0.045
mm
6.67

Expansion
(mm)
1

Water
Requirement%
88.9

Fly ash:Fly ash used in this study was manufacture by Shandong
Zouxian-China.the specific gravity of the fly ash is 2.4, loss
on ignition 0.48, the other properties of fly ash are presented
in Table 5. ASTM C618; the requirement for Class F andClass C fly ashes, and the raw or calcined natural pozzolans,
Class N, for use in concrete. Fly ash properties may varyconsiderably in different areas and from different sources
within the same area. The preferred fly ashes for use in highstrength concrete have a loss on ignition not greater than 3
percent, have a high fineness, and come from a source with auniformity meeting ASTM C 618 requirements [12].
Table 5: Physical Properties of Fly Ash
Moisture
Content %0.67

Loss onignition %2.7

Fineness Retain
0.045 mm %0.23

Density
Kg/m³
2373

3.2 Proportioning, Mixing and Casting of Specimens
There is no empirical method available for proportioning
high strength concrete. The procedure to get the proportionsin this study is the approach that recommended the in ACI
211.4R-08[13], by starting with mixture proportion that has
been used successfully on other projects with similar
requirements. Given this starting point, trial mixtures weremade in the laboratory and under field conditions to verify
performance with actual project materials. Hundreds of trial
batches were performed in the laboratory and several
adjustments were carried out in order to identify the optimum
proportions. The final optimum and best trials used in theconstruction of the concrete dam are shown presented inTable 6. A concrete fixed mixer with capacity of 0.125
m3was used, the mixes from Table 6 were scaled down
depending on number of molds for different tests, and themixer was buttered by mixing amount of cement, sand withwater because it is difficult to recover all the mortar from themixer. The mortar adhering to the mixer after discharging is
intended to compensate for loss of mortar from the test batch.
The following steps were to mix each batch; all the mixing
ingredients, including the mixtures, were scaled down andweight out. The coarse and fine aggregates, cement and othercementitous materials were added to the mixer. The mixer
rotated for 2 minutes (dry mixing). Superplasticizer was
dispersed in about 2/3 of water before added to the mixer andstarted rotated the mixer again for 2 minutes. The mixer was
shut off about 1 minute to let the aggregate absorb some ofthe paste, the aggregates were approximately in saturated
surface dry condition (SSD) at the time the batch was
prepared. The aggregates were sprayed with water andcovered by burlaps for at least 24 hours.
3.3 Curing and Testing of Specimens
Two different curing methods were used in this study; lime
saturated-water curing and the curing compound methods.
534

The effect of different curing method was studied andcommented on. After mixing, a portion of the fresh concrete
was placed aside for plastic properties determination. Slumpof fresh concrete was measured according to ASTM C143.
Precautions were taken to keep the slump between 150-200
mm to obtain pumpable concrete for dam construction.
Concrete casting was performed according to ASTM C192.
Molds were covered to prevent loss of water from
evaporation. Specimens were kept for 24 hours in molds at atemperature of about 23 C in casting room, and then cured forthe specified time at approximately 23 C ± 2 C. The
specimens were tested in dry state for compressive, flexural,and splitting strengths tests in accordance with BS EN
12390-2:2000[12], ASTM C78 or ASTM C29, and ASTM
C496 respectively.

days to 28 days strength of 0.8 ~ 0.9 for high strength and 0.7~ 0.75for normal strength (ACI). The high strength gain withage in this study of 3 to 28 and 7 to 28 days are 0.5~ 0.64 and0.75 ~ 0.86 respectively.

4. RESULTS AND DESCUSSION

Fig. 6: Variation of Compressive Strength with Time forDifferent Grads
Fig. 7 shows the variation of the compressive with thewater-cementitious materials ratio. Non-linear regression
analysis performed for the tow parameters concluded that an
exponential formula strongly related the compressive
strength and the water-cementitious materials ratio. The
exponential formula is shown in the figure. Fig. 8 shows theeffect of san content on the compressive strength of HSC.
Non-linear regression analyses performed have shown that aquadratic formula strongly relates these two parameters.

4.1 Optimum Proportions
Table 6 presents the optimum mix proportions for thedifferent grades used in the dam construction project. From
the table it is clear that five different grades of high strength
concrete (60, 70, 80, 90, 100MPa) were successfully
produced using local Sudanese aggregates and fly ash. Three
w/c ratios 0.22, 0.32, and 0.35 were found to produce themaximum values of strength in the different grades ofconcrete. SF and FA replacements in the range of 10 to 15%
for each one were found in the optimum combinations ofingredients to produce high strength concrete. Cement
content between 390 and 560 Kg/m3 for the five grades.

Table 6: Optimum Mix Proportion for Different Grades of HSC
Mix
DesignConcrete
Class

W/C

Sand
%

Cement(Kg/m3)Water(Kg/m3)Fly ash(Kg/m3)

HSC1
HSC2
HSC3
HSC4
HSC5
HSC6

50 MPa60 MPa70 MPa80 MPa90 MPa100 MPa

0.46
0.35
0.32
0.3
0.22
0.22

42
40
38
38
36
34

390
419
486
500
585
560

179
163
173
167
143
154

0
47
0
0
0
70

4.2 Compressive Strength
Fig. 6 shows the variation of strength concrete for thedifferent grades with time. Starting form three days up to 28
days, it is clear that it was possible to produce high strength
concrete (up to 110 MPa) with stable and acceptable rate ofstrength gain. Considering the type of cement used in thisstudy (ordinary Portland cement) and the relatively small
ration of SF, it is also clear that it was possible to produce
high strength concrete with very high rates of early strength
using the same local materials. Different relations wereobtained from data accumulated from the tests results. Highstrength concrete shows a higher rate of strength gain withage than normal strength. Parrott reported typical ration of 7

SilicaFume(Kg/m3)
0
0
54
56
65
70

Admixture(Kg/m3)
3.21
3.72
6.48
8.896
10.4
12.6

CrushedSand(Kg/m3)
798
746
667
665
621
522

Aggregate(5-20)(Kg/m3)
1102
1119
1089
1086
1104
1014

Unit
Weight(Kg/m3)
2473
2496
2475
2483
2528
2403

Fig. 7: Effect of Water-Cementitious Ratio on
Comperssive Strengh535

4.4 Tensile Splitting Strength
Fig.11 shows the non-linear exponential relation between the
splitting tensile strength and the compressive strength of
HSC. A non-linear quadratic relation was found to describe
the effect of water-cementitious materials ratio on the
splitting tensile strength. This relation is shown presented inFig. 12.

Fig. 8: Effect of Sand Content on Compersive Strength
4.3 Flexural Strength
Fig. 9 shows an exponential correlation between flexuralstrength and compressive strength of HSC. This correlation
obtained from the non-linear regression analysis performedfor data collected during the experimental program. A simple
linear relationship was found between the flexural strengthand the water-cementitious materials ratio and shownpresented in Fig. 10.

Fig. 11: Variation of Spliting strength with Compressive
Strength

Fig. 12: Variation of Spliting strength with
Water-Cementitious Ratio

Fig. 9: Variation of Flexural strength with Compressive
Strength

5. CONCLUSIONS

Fig. 10: Variation of Flexural strength with
Water-Cementitious Ratio

Based on the findings of this study the following conclusions
were made:i. When carefully mixed and cured, locally produced
concrete materials (aggregates, fly ash and ordinary
Portland cement) of good quality and with theiroptimum proportioning can be successfully used withother chemical and mineral admixtures to produce highstrength concrete of excellent properties.
ii. It is essential to optimize the doses of mineral andchemical at admixtures (silica fume, fly ash, and supper
plasticizer) when producing mass high strength concretein order to control the cost of materials.
iii. With other mix proportioning parameters held constant,the results of the present investigation indicated that the
maximum compressive and flexural strength occurred at536

about 10 to 15% Silica fume content and 20% fly ash
content.
iv. The present study shows that the maximum values ofcompressive strength for different grades were obtained
at water-cementitious materials ratios between 0.22 and
0.35.
v. Both compressive and flexural strengths of concrete
continue to increase as the percentage of silica fumeincreases.
vi. Non-linear regression analysis was used to correlatedifferent parameters involved in the characterization ofhigh strength concrete produced from local Sudaneseaggregates, such as compressive, tensile, and flexuralstrengths, water-cementitious ratio, sand content, andcementitious content.
ACKOWLEDGEMENTS
The authors thank SMEC International Company for theircooperation during the part of the experimental testing
program of this research. The authors also thank the Dams
Implementation Unit (DIU) of Sudan for allowing theauthors to use the information and photographs needed inthis research paper.
REFERECES[1] Berntsson, L.; Chandra, S. and Kutti, T. (1990),“Principles and Factors Influencing High-Strength
Concrete Production,” Concrete International,
December, pp.59-62.
[2] Cetin, A. and Carrasquillo, R. L. 1998. HighPerformance Concrete: Influence of Coarse
Aggregates on Mechanical Properties, Journal of
ACI Materials, 95:252-261.
[3] Muhammad A. Rashid & Mohammad Abul Mansur(2009), “Considerations in the Producing HighStrength Concrete”, Journal of Civil Engineering(IEB), 37(1) (2009) 53-63.
[4] ACI 363R-92: “State-of-the- Art Report on HighStrength Concrete”, American Concrete Institute,
Farmington Hills, Mich., 394 pp.
[5] ASTM C-150 Standard Specifications for Ordinary
Portland Cement, Type I. American Society for
Testing and Materials, Tests Specimens andLaboratory, Philadelphia, PA, 2003.
[6] ASTM C-136, Standard Test Method for Sieve
Analysis of Fine and Coarse Aggregates,
Philadelphia, PA: American Society for Testing andMaterials, 2001.
[7] ASTM C-33, Standard Specification for Concrete
Aggregates, Philadelphia, PA: American Society
for Testing and Materials, 2003.
[8] ASTM C-127 Standard test method for
Determination of The Specific Gravity andAbsorption of the Coarse Aggregates, Philadelphia,

537

PA: American Society for Testing and Materials,
2000.
[9] ASTM C-128 Standard test method for
Determination of The Specific Gravity andAbsorptionof the Fine Aggregates, Philadelphia,
PA: American Society for Testing and Materials,
2000.
[10] ASTM C-40, Standard test method for organic
impurities in sand for concrete, Philadelphia, PA:
American Society for Testing and Materials, 2003.
[11] ASTM C-494: Standard Specifications forConcreteSuper
plasticizing
Admixtures,
Philadelphia, PA: American Society for Testingand Materials, 2003.
[12] ASTM C-618:” Standard Test Method for Physical
Properties of Fly Ash”, PA: American Society for
Testing and Materials, 2001.
[13] ACI 211.4R-93:“Guide for Selecting Proportionsfor High-Strength Concrete”, American Concrete
Institute, Report 211.4R-93, Detroit, 1993.
[14] British Standards Institution, “Methods for making
concrete cubes in laboratory,” BS-1881-116: Part
110, 1990.
[15] British Standards Institution, “Methods for
Determination of compressive strength of concrete
cubes” BS 1881-108: Part 110, 1990.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

BEHAVIOR OF FOLDED PLATE ROOFING SYSTEMS WITH OPENINGSLOCATIONS*HALA ELKADY -**AHMED HASSAN*Professor, National Research Center** Beni Suef University, Egypt and Umm Al-Qura Univeristy, Saudi
Arabian
ABSTRACTThis paper investigates the effect of openings locations on the structural behavior of quadratic folded plate (Q.F.P) roofs
by using different geometric configurations for the main elements of the system. Furthermore, the impact of suchvariance on the behavior of the structure system under both static and dynamic conditions is investigated in this study.
The selected and investigated parameters in this study are openings locations and folded plates rise (height). Different
spans of 14, 20 and 26 meters were selected for all the investigated parameters. 3-D Finite Element Model (FEM) was
adopted to examine the suggested variables to meet the target of this study. Linear static analysis was performed toanalyze the effect of the investigated parameters on system deflections, moments, tension and compression stress.
Quadratic folded plate Q.F.P slabs with rise varying from 90 to 180 cm were studied. Results indicated that the differencein the rise reduced the roof deflection by 72%. Moreover, the behavior of the folded plate with openings at different
locations was improved.
Also, results indicated that the maximum bending moment for intermediate beams increased by 69% in case of center
openings locations at beam center. The maximum bending moment at 0.57 L (intermediate beam length) for quarter
openings locations was increased by 64%. The edge openings locations had a slight effect on the diaphragm bending
moment and the intermediate beam fixed end moment.
Three Dimensional dynamic modal analyses were performed and the effect of different openings locations on thefundamental modes was investigated. Results of modal analysis showed that openings locations had no significant effecton the fundamental frequencies and Fundamental mode shapes. The obtained results from this study emphasized on theimportance of elaborate numerical analysis to deal with such sensitive models, as the geometric properties and openings
locations for each contributing element which had a clear effect on the overall performance of the system.
Finally, model results proved that folded plate openings locations at center is the most effective parameter compared tothe others investigated parameters, while edge openings locations in the folded plate had the lowest effect on both staticand dynamic behavior of the investigated system.
Keywords: Folded plates; Height; Openings location; Free vibrations

1. INTRODUCTION
Folded plate is one of the most practical structures used incivil engineering applications since it has many merits such as
lightweight, easy to form, low cost and high resistance forloads. It is used in many applications such as roofs, sandwich
plate cores and cooling towers. As the use of folded plates has
many merits compared to flat plates, it is essential toinvestigate their behavior with openings and without
openings under different load conditions. Subsequently, there
are several methods available to analysis this type of structure(1 to 8). Conventional analysis methods are simple and easy
while they have some limitations of generality of applicationand precision.
Early researchers solved folded plate problems
approximately with the use of beam method or the theory
that neglected the relative joint displacement (2). However,these two methods met difficulties in dealing with thegeneralized folded plate problems. Both computational
approaches and numerical methods for the analysis of thefolded plates offer more precise solutions compared toConventional analysis methods. The methods of interest areincluding finite strip methods, one of the earliest work
presented based on this method was introduced by Cheung(1969) through the work presented by Golley and Grice(1989) along with the work presented in the same year by
Eterovic and Godoy (1989). The combined boundary
element-transfer matrix method [Ohga et al., 1991] andfinite element method (FEM) which was a topic of research
for many previous researchers [Liu and Huang, 1992; Perryet al., 1992; Duan, 2002]. From these methods, the FEM is
the most convenient method because it can be applied toanalyze large complex structures as well as all kinds ofboundary conditions and loadings can be easily
implemented. Subsequently, most of commercial software
used for structural analysis used FEM method.
Quadratic folded plate Q.F.P roofs are one of the mostcommon types used for roofing systems and that is why it
was selected in this study. The presented parametric study
on the effect of openings locations on folded plate behavior
were examined by using Finite Elements analysis and Linear
Static Three Dimensional F.E. analysis. The effects of
openings locations and folded plate rise on slab deflection,intermediate beam moments, and stress distribution on slab,

538

maximum stress were executed.
In some cases, Quadratic folded plate Q.F.P slabs systemshould include openings. They may have small dimensions,like those needed to accommodate heating, plumbing, andventilating risers, floor and roof drains, and access hatches
but in some cases openings may be larger than the code
limitation in size and position. The large openings presencein slab system reduces the stiffness and increases thedeflection and then there are previous studies investigatedthe openings on the behavior of flat plate structure.
Based on the output of models, design charts for thisparticular system will be presented to select the best openings
locations and also the parameters effect on the behavior of thestructural system will be evaluated. Recommendations will
be given in accordance with the impact of openings locations
on the behavior of structural system.
2. FINITE ELEMENTS ANALYSIS
2.1 Geometry and Dimensions of the Investigated Systems
Q.F.P. panels of 7.8 ms width were studied as presented in
Figure 1. This width was kept constant throughout theanalysis, while the three different spans of 14, 20, and 26 m
were changed to investigate their effect on the behavior ofstructural system. A constant openings size of (3m×2m)
which is located in the top surface of the folded plate slab at
different positions (center- quarter – edge) were considered
for each panels. Models heights are varying from 90 to 180
cm. A 3-D F.E. analysis was performed and the roof wasmodeled using 3-D quadratic shell elements which arepresented in Figure 2. While intermediate beams, end
diaphragms, and columns were modeled as 3-D frame
elements as shown in Figure 2.

Figure 2: 3-D model of investigated folded plates panels
and center openings locations.
2.2 Static Analysis of Q.F.P System
3-D Static linear F.E. analysis was performed by applying
each of the parameters under study separately as openings
locations and folded plate rise. An imposed static load of 150
kg/cm2 was applied on all the models, as a make up for
flooring and service load. Throughout the analysis, thefollowing straining actions and deformations are monitored
and compared. Deflections and stresses of the folded plate arechecked and the maximum bending moments at theintermediate beams and diaphragm are reported. The effect of
the tested parameters will be presented in details later.
2.2.1 Effect of Openings Locations on the Q.F.P. System
Maximum Deflection for the Constant Rise

Figure 1: Cross section dimensions of investigated models

The first investigated parameter was the effect of the folded
plate openings locations on the maximum deflection at thecenter of the folded plate. There are three different spans of
14, 20, and 26 meters were studied. Figure 3 shows the
maximum deflection for the folded plate slabs at the center
point in the case of different lengths used versus openings
locations for folded plate. In case of center openings
locations, it is obvious that the maximum deflections for the
span length of 26 m and 0.9 m rise is increased from 47.6 mm
to 58.2 mm, with a percentage increase of 22%. The increaseof the central deflection in the case of quarter openings
locations was found 51.8 mm, with a percentage increase of 9%. The central deflection was found 47.6 mm in case of edge
openings locations, which has minor, effect on the center
deflection point. The measured deflection value in these cases
is approximately the same as in the case of control model.
For the 20 m Q.F.P span when the rise of folded plate equals
0.9 m, the central deflection was found 15.7 mm as presented
in Figure 3. The openings locations have a significant effecton the increase of the central deflection in the mostinvestigated cases. For example, the deflection increased to19.3 mm and 17 mm, which equals 23% and 8% in case of539

center and quarter openings locations respectively, in someinvestigated points compared to the control case.
The central deflection for the folded plate with 14 m span was
found 3.73 mm and this case is considered as a control model.
The effect of the openings locations on the central deflection
were 4.45 mm (increasing by 19 %), 3.89 mm (increasing by
4 %), and 3.77 mm which corresponding to center, quarter,and edge openings locations, respectively.

moment was found 28.37 meter ton (increasing by 67 %). For
the 20 m and 14 m folded plate lengths, the maximum
bending moments increased by 60% and 42 % respectively.
The effect of opening locations at quarter or edge has noeffect on the maximum moment at the center and this resultclose to the control model result as shown in (Fig.6). The
maximum moment was found at the quarter of theintermediate beam at the openings locations and it increasedby 64 %, 46% and 30% which corresponding to different
model lengths of 26 m, 20 m and 14m respectively as shownin Figure 5. The edge openings locations have a limited
significant effect on the intermediate beam bending moment
as shown in Figure 4.
Figure 4-6 gathers the effect of both opening locations and
Q.F.P. rise on the bending moment of the intermediate beam.
It is obvious from these Figures that for all investigatedmodels, the increase of the folded plate rise from 0.9 to1.8 m
results in a reduction for moment of the intermediate beam at
different locations.

Figure 3: Q.F.P. slabs maximum deflection versus
different opening locations for different rise.
2.2.2 Effect of Folded Plate Rise to the Maximum
Deflection
The investigated folded plate rise in this study was varying
from 90 cm to 180 cm. The folded plate rise has a significant
effect on decreasing the central deflection as shown in Figure
3. For the longest investigated span of 26m, the increase of
the folded plate depth (rise) from 0.9 m to 1.8 m with center
openings locations results in the reduction of deflection from58.2 mm to 16.5 mm. While, the deflection induced at thequarter openings locations reduced from 51.8 mm to 14.8 mm
in the case of the folded plate rise 0.9 m to 1.8 m.
The case of 20 m model doubling, the folded plate rise from
0.9 m to 1.8m reduced the central deflection from 19.3 mm to5.62 mm at the center openings locations while at the quarter
openings locations reduced the center point deflection from
17 mm to 5 mm.
In the case of 14 m folded plate span, the reduction percent on
the central deflection was found 65 % at both center andquarter opening locations as presented in Figure 3. The Q.F.P
slab rise is reduced the effect of openings locations on the
maximum deflection.

Figure 4: intermediate beam fixed end moment versus
different opening locations.

2.2.3 Effect of Folded Plate Openings Locations on theIntermediate Beam Moments.
Three trials for the openings locations of Q.F.P. slabs wereconsidered in this study; center, quarter and edge. These
openings have a significant effect on the increase of the
maximum bending moment of the intermediate beam and thevalue of moment is depending on the opening positions. In thecase of 26 m length without openings, the maximum bending
moment at the center of the intermediate beam was found16.8 m.t, while at the center openings locations the bending
540

Figure 5: intermediate beam moment at quarter versus
different openings locations.

Q.F.P.slab second mode shape
Figure 6: maximum moment of intermediate beam versus
different openings locations.
2.3 FREE VIBRATIONS ANALYSISThe investigated systems were subjected to free vibrations
analysis. Eigen values analysis was adopted in the linear
dynamic analysis. For the investigated spans, with different
parameters arrangement of mode shapes was consistent.
Columns sway dominates the first three modes and then theroof deformations is controlled the next ones. Figure 7
displays the roof modes and it can be seen from this Figurethat the dominating mode for the Q.F.P. slab is intermediate
symmetric bending mode. The second mode wasanti-symmetric bending, followed by symmetric bending of
the external spans and then followed by alternative modes ofanti-symmetric and symmetric plate bending.
In Figure 8, the fundamental frequencies of the system were
given for Q.F.P. slabs with spans of 14, 20 and 26 meters. It isnoticed from this Figure that the first three modes werecolumns lateral displacements. Fundamental modes for thefolded plate roof started from the fourth mode. The frequencyof three fundamental modes for the roof of 14 and 20 metersspans were found very close, while the 26 meters frequencies
was found 27% lower than the 14, and 20 meters spans.
Figure 9 presents the effect of the roof rise on the fundamental
frequencies of the system with includes span of 20 meters.
Two rises were investigated which were 90, and 180 cm. It isobserved that the system with higher rise of (180 cm) has
higher fundamental frequencies by 20%. This result is
probably attributed to the higher stiffness of the used roof.
The change in the openings locations has no effect on the roof
modes as well as the fundamental frequencies when freevibration analysis is subjected.

Q.F.P.slab third mode shape

Q.F.P.slab fourth mode shape

Q.F.P.slab fifth mode shape
Q.F.P.slab First mode shape541

Figure 8: Fundamental modes frequencies for folded
plate roofs with different spans.
Q.F.P. slab with edge opening third mode shape

Q.F.P.slab with edge opening fourth mode shape

Figure 9: Effect of folded plate rise on fundamental
frequencies under free vibrations.
3. DISCUSSION OF RESULTS

Q.F.P.slab with edge opening fifth mode shape
Figure 7: Fundamental mode shapes of Q.F.P.slab

For the tested models edge openings locations has a limited
effect on the Q.F.P roofs central deflection, maximum
bending moment, edge moment of intermediate beam and thediaphragm bending moment. This effect did not exceed morethan 3% as shown in Figure 10. The center openings locations
have a clear effect on the central deflection compared to thedefections induced in both quarter and edge openings
locations. The center openings locations increased deflection
by 14 % more than quarter opening effect for 14 m and 20 m
but in Case of 26 m this percentage was 12%.
Doubling the height of the investigated folded plate from 0.9
meters to 1.8 meters is associated with a reduction in
deflection by an average value of 72% as shown in Figure 12.
From the same figure it is shown that there is no variation in
the increase of the percentage of central deflection for thedifferent spans used from 26 m to 20 m, while the effect of the
opening locations at 14 m folded plate length was foundlower than the induced in the cases of 20 m and 26 m folded
plate spans.

542

moment at intermediate beam center and edge moment
observed.

Figure 10: Q.F.P. slabs maximum deflection versus
different opening locations for different length.
Figure 12: Intermediate beam moment versus different
openings locations.

The variations of openings locations have minor effect on the
maximum tensile stress. This effect do not exceed than 4%.
These openings have a great effect on the compression stress,
which increased by 174%, 170%, 120% for the different
spans of 26 m, 20m and 14m respectively. The quarter
openings locations increased the compression stress by 56%,
30%, and 5% for different spans 26 m, 20m, and 14m
respectively.
The openings locations from center to edge have a limited
effect on the fundamental frequency of the Q.F.P slab. Modeshapes patterns arrangement was not affected by the variationin the openings locations from center to the edge. In otherwords, it can say that the mode shapes came in the same order
for all the investigated spans.

Figure 11: Q.F.P. slabs maximum deflection versus
different opening locations for different length.
The maximum moment location was found at theintermediate beam center in the case of the control model andcenter openings locations model. In the case of quarter
openings locations, the maximum moment's position was
found at a length of 0.75(at openings locations) for theintermediate beam length. The maximum moment in case ofedge openings locations was found close to the maximum
moment of control model as presented in Figure 12.
Varying the openings locations has a significant effect on the
maximum moment of intermediate beam. The center
openings locations increased the maximum moment by 67%,
60% and 42% which corresponding spans of 26 m, 20 m, and
14 m, respectively. The maximum moment is located at thequarter of the intermediate beam at openings locations and it
increased by 64 %, 46% and 30% which corresponding todifferent model lengths of 26 m, 20 m and 14m, respectively
as reported in Figure 12. The openings locations have alimited effect on the edge moment and diaphragm bending
moment, which did not exceed more than 3%. At the quarter
openings locations there is no effect on the maximum

4. CONCLUSION

543• The propose approximate static solutions for the fixedboundary conditions at the end diaphragms for the
Q.F.P slab ends (9, 10, and 11) are not reliable forthe longer spans, especially for spans higher than 20
meters. Subsequently, the stiffness should berecommended in the numerical modeling.
• Increasing folded plate rise was the most significant
factor in improvement the static analysis for
deflection control, on the other hand increasing theroof rise lead to an increase in the fundamental
frequencies of the system.
• Rise effect was most effective in shorter spans.
• Increasing folded plate rise enhanced the structural
behavior of the overall Q.F.P. system.
• Edge openings locations have a limited effect on the
Q.F.P roofs static and dynamic behavior.
• Center and quarter openings locations have an effecton the change in static deflections, or straining
actions of the system.

• Spans from 14 to 20 meters for the investigated Q.F.P.
slabs had very close fundamental frequencies, thiseffect widens noticeably on analyzing longer spans.
• Center openings locations has a great effect on the
Q.F.P. slabs deflection greater than quarter openings
locations by 14%.
• The variation of the openings locations has asignificant effect on the maximum moment ofintermediate beam, while the center openings
locations increased the maximum moment by 67%,
60% and 42% which corresponding spans of 26 m,
20 m, 14 m, respectively.
• The openings locations has a limited effect on theedge moment and diaphragm bending moment,
which did not exceed more than 5%• At the quarter openings locations there is no effect on
the maximum moment at intermediate beam centerand edge moment observed. The maximum moment
is located at the quarter of the intermediate beam at
openings locations and it increased by 64 %, 46%
and 30% which corresponding to different model
lengths of 26 m , 20 m and 14m, respectively.
• The openings locations from center to edge have alimited effect on the fundamental frequency of the
Q.F.P slab.
• Mode shapes patterns arrangement was not affectedby the variation in the openings locations from
center to the edge. In other words, it can say that themode shapes came in the same order for all theinvestigated spans.
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[14] Bernadou .M, “Finite Element Methods for Thin
Shell Problems”, J. Wiley & Sons, 1996.
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considerations for the finite element analysis of shellstructures”, Comput. Struc. 66 (1998) 19–36.
[16] Liew KM, Wu YC, Zou GP, Ng TY.,
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Comput Methods Appl Mech Eng

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Improving concrete compressive strength using natural additive
Abdoullah Namdar, Ideris Bin Zakaria and Nurmunira Binti Muhammad @ Atan
Faculty of Civil Engineering & Earth Resources
Universiti Malaysia Pahang, Malaysia
anamdar@ump.edu.my
ABSTRACT
Construction material quality is one of the important issues for eliminating any types of problems may face civilengineering structures. For improving concrete quality and reducing construction cost a simple procedure has been
evaluated. Thermally treated bentonite of 6% quantity of cement content in concrete mixture design is used as aconcrete additive to assess concrete compressive strength. The XRD experimental for studying crystal structure ofmodified bentonite has been conducted. The result indicated acceptable improving concrete compressive strength.
And in compare to the natural additive cost the method has economic effective, and can be produced at any
situation, and helps to reduce construction cost especially for low cost construction.
Keyword: Bentonite; crystal structure; compressive strength; natural additive; low cost construction
1. Introduction
The mineral additives could have different effects on theproperties of concrete due to different mineralogical andchemical process [1]. The mineral admixture has
different effect on high strength concrete. And full loads
apply to concrete structure on early age is important to a
construction programme [2]. There is an investigation on
natural mixed minerals to analyze compressive concrete
strength through effect of natural mixed minerals on
composition of the gel phase and its structure [3]. It has
been instructed that the mineral admixture in concrete
mixed design decreased both unit weight andcompressive strength [4]. From this point of view
European Standards (EN 206-1) restrict the use ofmineral additions to low proportions during the
manufacture of concrete [5]. It is due to some mineral
addition in high percentages as a replacement for cement
can result in a concrete with lower compressive strength.
From other hand using proper combination of minerals
results in produce economical concrete and CO 2
emissions reductions for sustainable development [6].
To enhancement of concrete strength using mineral
mixture technique some empirical relations has been
investigated [7]. The rational methods have also been
proposed to analysis mineral additives effect on concrete
strength in specific age [8]. There is reported an artiﬁcial
neural networks study for predicting the corecompressive strength of self-compacting concrete (SCC)
mixtures with mineral additives [9]. In view ofsustainable development, mineral admixture commonlyis used to replace part of cement in concrete as
supplementary cementitious materials [10], [11]. One of
method is performance of steam cured concrete byadding mineral admixtures to determined concrete
compressive strength in short and long term tests [12].

The mineral admixtures to partially replace cement
preserve the non-renewable resources required for the
production of cement [13-15]. In attention to the mineraladmixture is utilized for many years either as
supplementary cementitious materials in Portland
cement concretes or as a component in blended cement[16]-[20], the objective of this study is to asses best heatlevel in modify bentonite to develop appropriate mineral
under laboratory condition for enhancement concrete
compressive strength in early age, produce low costconcrete, reducing chemical additive and cement
consumption in concrete mixture design which decreases
environment pollution.
2. Methodology and Experiments
In this research work the bentonite has been selected for
improve concrete compressive strength. The bentonite
was subjected to the heat for 1 hour for 200 ºC, 400 ºC,
600 ºC and 800 ºC. The concrete mixed design is madeto reach acceptable compressive strength. The unheatedand heated bentonite hais mixed in concrete mixture
design in 6% quantity of consumed cement. The
concrete compressive strength for 7 days, 14 days and 28
days have been tested. To study macro properties ofbentonite-cement mixture X-ray is used.
3. Result and discussion
In an experimental investigation on Pakistani bentonite(from Jehangira, Swabi District) is treated by heat at 500°C and 900 °C and the cement in mortar and concrete
was replaced with the bentonite at 0%, 20%, 30%, 40%
and 50% by cement mass. And used as a cement
replacement material in mortar and concrete, it wasreported that Pakistani bentonite can be used to replace
up to 30% of cement to produce concrete with sufficient

545

compressive strength for low-cost construction resistantto sulphate attack [21]. The proportion of bentonite,(3%, 6%, 9%, 12%, 15%, 18% and 21% by weight of
cement) is used as replacement. And has been foundthat the workability, fresh concrete density and water
absorption decreased as bentonite usage is increased.
The comparative compressive strength analysis indicatedthat at 3 days of testing, the mixes containing bentonite
showed lower strength than control mix while at 56 daysof testing, the bentonite mixes showed higher strengththan the control mix [22]. When bentonite is thermallytreated by heat for one hour under 200 ºC theMontmorillonite-15A has been developed, this mineral
has strong negative effect on concrete compressive
strength (table 1 and figure 1). In second part ofexperiment in thermally treated of bentonite, 400 ºC,heat for one hour has been used to observe newmodification in bentonite. In this stage Rectorite is
created. The Rectorite mineral has reduced concrete
compressive strength significantly. Up to this level ofresearch thermally treating bentonite is presented
disappointing results. The 600 ºC heat is submitted to thebentonite for one hour, several minerals have beendeveloped (figure 2 and table 2), these minerals improve
concrete compressive strength due to number ofminerals it is difficult to conclude effect of them on
improving concrete compressive strength. In continuing
investigation 800 ºC heat has been applied to thebentonite for one hour and quantity of 6 % newdeveloped minerals is added to concrete mixed design.
The concrete compressive strength test had shown that in
improving around 8% concrete compressive strength inearly age of 7 days and also 13% in 28 days.

Table. 1. Concrete compressive strength
7days
14 days% ofConcrete typeStressStressBentonite(N/mm2) (N/mm2)
Unmixed
25.77
30.86
bentonite
Unheated
6
23.13
27.76
bentonitebentonite 200 ºC
6
15.5
18.46
bentonite 400 ºC
6
11.8
14.71
bentonite
6
27
30.36
600 ºC
bentonite
6
27.95
35.65
800 ºC

28 days
Stress(N/mm2)
34.42
31.24
27.77
23.33
31.5
38.95

Table. 2. Minerals is developed in bentonite thermally treatedBentonite type
Name of minerals
Unheatedalpha-Si O2, quartz low high HP, synMontmorillonite-15A,
200 ºCalpha-Si O2, quartz low,
Rectorite,
400 ºCalpha-Si O2, quartz low,
Zinc aluminum succinate dihydrate,
3-Amino-4-nitro-6-methyl-8-oxopyrazolo(1,5a)pyrimidine,
600 ºC
Sodium Aluminum Silicon Oxide,
Zircon,Quartz alpha, alpha-Si O2,
Cobalt Sulfide,
cyclo undecasulfur,
800 ºC
Quartz, syn,546

Fig. 1. Concrete compressive strength in 7, 14 and 28 days

Fig. 2. XRD pattern of thermally modifying bentonite
5. Conclusion
The mineral mixture in concrete mixed design has
significant effect on setting time and final strength inconcrete. The quantity of used thermally treatedbentonite is very less and it made this product economic
and also helps in reducing CO 2 producing in theenvironment.

547

Rectorite and Montmorillonite-15A have strong negative
effect on concrete compressive strength. The low costconcrete with better compressive strength is main
product of this research work.

References
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2007, pp. 923-37.
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Hassani, A.A. Ramezanianpour, “Effect ofadding mineral additives to alkali-activatednatural pozzolan paste,” Construction and
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performances, production et conformité,
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concretewithhighvolumemineraladmixtures,” J Build Environ, vol 42, 2007, pp.
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incorporating ﬂy ash or a chemical admixture,”
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conference on the use of ﬂy ash, silica fume,slag and other mineral by-products in concrete,
Canada, SP-79, 1983, pp. 333-46.
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fume and ﬂy ash,” Oslo: Nordic Concrete
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depth of concrete with ﬂy ash,” In: Proceedingsof the third international conference on ﬂy ash,silica fume, slag and natural pozzolans in
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vol 30, 2012, pp. 237-242.

DEM Simulations and Laboratory Experiments on Physical and Mechanical
Properties of Sand-Gravel Mixtures
Janaka J. Kumara1, Kimitoshi Hayano2, Yuuki Shigekuni2 and Kota Sasaki2
1 Dept. of Civil Engineering, 2 Dept. of Urban Innovation, Yokohama National University, Japan
ABSTRACTIn railway tracks, ballast fouling due to fine materials mixing with ballast has been identified as a challenging issue
nowadays. In this research, deformation characteristics of sand-gravel mixtures, repressing fine-ballast mixtures, werestudied using laboratory and DEM simulated triaxial compression tests. DEM simulations were done in Yade, an open
source developed based on DEM. Initially, void ratio characteristics of sand-gravel mixtures were studied. After void
ratio results were obtained, triaxial compression tests were conducted on specimens with 50% and 80% of relative
densities. Void ratio results indicated that the specimen with 30% sand showed the smallest void ratios (i.e., the densest
packing) in DEM simulations. Triaxial test results indicated that the specimens with 30% sand experienced the highest
deviator stress in both 50% and 80% of relative density cases, showing clear relation of void ratio and deviator stress.
Keywords: DEM Simulations, Laboratory Tests, Mechanical Properties, Physical Properties, Sand-Gravel Mixtures

1. INTRODUCTION
In railway tracks, due to repeated heavy train loads, finematerials mix with ballasts. The fine materials come mainly
from the underneath layers and to a lesser extent due toparticle crushing as well [1]. Changes in deformation
properties of ballasts due to sand intrusions have been
identified as a major problem in railway engineering. In
many cases, ballasts containing sands shows large
settlements, which is induced mainly by railway trafficloading. However, degree of settlement depends not only
on traffic volume but also on physical and mechanicalproperties of the ballasts. Though there had been various
researches conducted on pure materials of ballasts and fines[2]-[3], there had not been sufficient researches conducted
on deformation characteristics of fine-ballast mixtures
simulating actual field rail-track conditions. Perhaps,difficulties in simulating multi-size particles in numericalsimulations would have been the main issue behind lessnumber of researches on fine-ballast mixtures. However, it
is very important to understand effects of fine materials
mixing on deformation characteristics of ballast tounderstand degradation of ballast layers and to propose
maintenance works for degraded ballast layers.
After development of DEM [4], DEM simulations
became the most widely used numerical method to study
deformation characteristics of granular materials. As
laboratory experiments are complicated and expensive forvarious types of field conditions, nowadays, numericalsimulations are preferred in many research works.
However, laboratory experiments are still required to verify
the accuracy of the numerical simulations. In this research,deformation characteristics of granular materials including
sand-gravel mixtures were studied using triaxial
compression tests in Yade [5] as the numerical method in
addition to laboratory triaxial compression tests.

1.1 Yade
Yade is an extensible open-source framework for discrete
numerical models, focused on Discrete Element Method.
Yade is a 3-D numerical method. The original code wasmodified to obtain particle size distribution of sand-gravel
mixture since the original code does not allow gap gradedparticle size distribution curves like in sand-gravel
mixtures. Widulinski et al. (2009) [6] and Sayeed et al.
(2011) [7] have also conducted triaxial compression tests
using sphere particles in Yade recently. They found good
agreement with experimental results.
2. METHODOLOGY
In this research, since size of ballast (i.e., 10-60mm) is toolarge to use in triaxial compression test (specimen size is
100x200mm), gravel with the size of 1/5 of actual ballast,
was selected. Medium size sand was used as the finematerial. At the initial stage, void ratio characteristics ofsand-gravel mixtures were evaluated using laboratory
density tests and DEM triaxial simulations. Density tests
were conducted according to JIS A 1204-2009 [8]. In thenext stage, deformation characteristics of sand-gravel
mixtures were evaluated on the specimens prepared with
50% and 80% of relative densities. Deformation
characteristics were evaluated using both laboratory and
DEM simulated triaxial compression tests.
2.1 Laboratory Triaxial Compression Tests
Laboratory triaxial compression tests were conductedaccording to JGS 0527 [9]. Triaxial compression tests wereconducted using the apparatus shown in Fig. 1. Axial
deformations were measured by an external displacementtransducer and a pair of LDTs [10]. However, themeasurements of LDTs used to determine small strain
deformation properties, are not included in this paper.
Triaxial compression tests were conducted under 80kPa ofconfining pressure.

549

Dr =

e max − e× 100(% )e max − e min(1)

2.3 Sample Preparation
Six specimens each were prepared with different amount offines as given in Table 2 for both laboratory experiments
and DEM simulations. Two specimens were prepared withsand and gravel separately to compare the results with thoseof sand-gravel specimens.

Fig. 1: Triaxial apparatus
2.2 DEM Simulations
In triaxial test simulations, void ratio is determined by
friction angle, φ c , during isotropic compression. Aftermaximum and minimum void ratios, e max and e min , of thespecimens were determined, void ratios related to 50% and
80% of relative densities, e 50 and e 80 , were obtained using
Eq. (1). Then, the values of φ c related to e 50 and e 80 were
determined. In DEM simulations, particle size wassimulated as 100 times larger than the size of laboratory
specimens to reduce simulation time. DEM simulations
were done using sphere particles. The input parameters
used in DEM simulations are given in Table 1. Though
5m/s of V w was used for gravel specimens, smaller valuesof V w (1m/s used here) were required for the specimens ofsmall size particles to stabilize the simulations.

3. RESULTS AND DISCUSSIONS
Fig. 2 shows particle size distribution of laboratory
specimens. Particle size distribution of soil samples wasevaluated using sieve analysis test according to JIS A 1204
[11]. Fig. 3 shows particle size distribution of DEM
simulations. As shown in Fig. 3, particle size of DEM
simulated specimens is 100 times larger than that oflaboratory specimens.

Table 1: Input parameters of DEM triaxial simulations
Parameter
Value
Remarks
Help to stabilize
Stability criterion
0.01
systemFriction angle of spheres
See φ cdetermines
duringisotropic
Table 2 void ratio
compression, φ c
Coefficient of Cundal
0.2 Default value
non-viscous damping
0.000658 Default value
Max. time step, ∆t (s)Density of spheres,2600ρ s (kg/m3)
Max. velocity of walls,
Help to stabilize
1.0/5.0*
V w (m/s)
the system
Number of particles, N
10000
Confining pressure,80
σ c (kPa)
Friction angle of spheres
30
just before shearing, φ s(degree)
Ratio of shear and normal
contact stiffness for
0.5
spheres, K s /K n
Stiffness of spheres,15
E s (MPa)
Strain rate (s-1)
0.1Note: * Used for gravel specimen
550

100

Sands (%)
0
15
30
50
70
100

80

Percent passing (%)

Specimen

Table 2: Details of friction angle and void ratio
Friction angle,
Void ratio, eFines
φ c (degree)(%)
D r = 50% D r = 80% D r = 50% D r = 80%
0
16.5
4.5
0.632
0.566
15
11.8
3.8
0.485
0.429
30
16.3
4.2
0.403
0.342
50
20.0
5.0
0.450
0.396
70
17.15
4.4
0.519
0.460
100
16.4
3.5
0.646
0.574

60
40
20
0

1

2

3

4

5

Grain size, D (mm)6

7 8 9 10

Fig. 2: Particle size distribution of laboratory specimens
100

Sands (%)
0
15
30
50
70
100

80

Percent passing (%)

Load cell

60
40
20
0

1

2

3

4

5

Grain size, D (mm) x 102

6

7 8 9 10

Fig. 3: Particle size distribution of DEM simulated
specimens

The results also showed that void ratios of DEM simulated
specimens are smaller than laboratory specimens for the
specimens with larger particles (e.g., gravel specimen). It
should be, however, noted that void ratio of DEM simulated
specimens were measured at the end of isotropic
compression (under 80kPa) whereas in laboratory
specimens, void ratios were measured before the isotropic
compression. As shown in Fig. 4, the difference in void
ratios of laboratory and DEM simulated specimens become
small for the specimens with more sands than those withmore gravel particles. The large difference in void ratios forthe specimens with larger particles should be attribute todifferent particle shapes in laboratory specimens and DEM
simulations. Fig. 4 also shows that DEM simulations cangive same void ratios as laboratory results for sands (i.e.,
for specimens with small particles). Fig. 5 shows relation ofe max and e min for both laboratory and DEM simulated
specimens. Fig. 5 clearly shows that both relations arelinear with a same gradient (i.e., increasing rate) thoughexperimental data shows higher void ratios. As explained,
smaller void ratios in DEM simulations should be attributeto change in particle shapes in the two methods.
emax,exp
emin,exp
e50,exp
e80,exp

Void ratio, e

1.0
0.8

emax,DEM
emin,DEM
e50,DEM
e80,DEM

0.6
0.4

0

20

40

60

80

Experimental data
DEM data

1.0
0.9

emax,exp = 0.226 + 1.089emin,exp

0.8
0.7
0.6

emax,DEM = 0.130 + 1.176emin,DEM

0.5
0.3

0.4

0.5

0.6

0.7

Minimum void ratio, eminFig. 5: Comparison of relations of maximum and minimum
void ratios

3.2 Triaxial Test Results
Fig. 6 shows a DEM triaxial test simulation at different
steps. Specimens were prepared in a box as shown in Fig. 6.
As shown in Fig. 6 (a), at the beginning of particle
generation, packing is very loose. At the end of isotropic
compression (i.e., Fig. 6 (b)), void ratios were obtained tocompare with those of laboratory specimens. Fig 7 showsdeviator stress vs. axial strain for laboratory specimens of
50% of relative density. Fig. 8 shows deviator stress vs.
axial strain for DEM simulated specimens of 50% of
relative density. As shown in Fig. 7, the specimen with 30%
sands experienced the highest deviator stress, followed by
the specimen with 15% sands. The results also showed thatgravel specimen experienced higher deviator stress thanthat of sand specimen. In DEM simulations too, as shown in
Fig. 8, the specimen with 30% sands experienced thehighest deviator stress, followed by the specimen with 15%
sands. In DEM simulations too, gravel specimen
experienced higher deviator stress than that of sandspecimen. However, it should be noted that laboratory
specimens experienced higher deviator stress compared to
DEM simulated specimens. The difference in deviator
stresses between the two methods should be attribute todifferent particle shapes as DEM specimens consist of
sphere particles while laboratory specimens consist ofirregular shape particles. Lin and Ng (1997) [13] also
showed that irregular shape particles give higher devistor
stress than that of sphere particles. He compared deviator
stress for the samples made of ellipsoid and sphere particles.
Yan et al. (2009) [14] also studied effects of particle shapes
on strength characteristics and found that sphere particles
give smaller stress ratio compared to clump particles. He
also observed that DEM simulations (using PFC3D) givesmaller stress ratio than that of experimental specimens.

Sand

Gravel0.2

1.1

Maximum void ratio, emax

3.1 Void Ratio Characteristics
Fig. 4 shows relation of void ratio of the specimens with %sands. In Fig. 4, e max,exp and e max,DEM represent maximum
void ratios of experimental and DEM simulated specimens
respectively. In Fig. 4, e 50 and e 80 represent void ratios ofthe specimens with 50% and 80% of relative densities
respectively. As shown in Fig. 4, void ratios (both e max ande min ) reduce with sands at initial level, then, after certain
amount of sands, increase with sands. Minimum value ofvoid ratios (both e max and e min ) reached at 50% and 30% of
sands for laboratory and DEM simulated specimens
respectively. The difference in amount of sands to reachminimum value of void ratios would be attribute to particle
shape as DEM simulations were done using sphere particles
whereas laboratory specimens consist of irregular shapeparticles. Lade et al. (1998) [12] also showed similar results
for binary mixtures of sphere particles where minimum
values of void ratios were observed mainly at 20% - 40%
fines depending on diameter ratio of two mixtures.

100

Sands (%)
Fig. 4: Void ratio of laboratory and DEM simulated
specimens551

Deviator stress, q (kPa)

500

30% sands 15% sands

400
300

Gravel200

70% sands
Sand

50% sands

1000Dr = 50%
0

2

4

6

8

10

Axial strain, εa (%)

12

Fig. 7: Deviator stress vs. axial strain from laboratory
specimens with 50% of D ra140

30% sands

Deviator stress, q (kPa)

120

15% sands

GravelSand

100

70% sands
50% sands

80
60
40

Dr = 50%

20
0

0

2

4

6

8

10

12

Axial strain, εa (%)
Fig. 8: Deviator stress vs. axial strain from DEM
simulations with 50% of D rbcFig. 6: A DEM triaxial test simulation (a) at the particle
generation, (b) after isotropic compression and (c) at theend of simulation (a sample with 30% sands)

14

14

Fig. 9 shows deviator stress vs. axial strain for laboratory
specimens of 80% of relative density. Fig. 10 showsdeviator stress vs. axial strain for DEM simulated
specimens of 80% of relative density. As shown in Fig. 9,
some specimens (i.e., the specimens with 30%, 50% and
70% sands) reached a peak value in deviator stress and thenreduced deviators stress same as a dense specimen. Out ofthese three specimens, the specimen with 30% sands
experienced the highest deviator stress. However, gravel
specimen experienced the highest deviator stress showing
stress-strain behaviour of a loose specimen where there isno any clear peak value in deviator stress. As shown in Fig.
10 for DEM simulations, the specimen with 30% sands
experienced the highest deviator stress, followed by the
specimens with 50% and 70% sands until 4% axial strain.
However, after 4% axial strain, gravel specimen
experienced higher deviator stress than the specimens of
50% and 70% sands. In 80% of relative density case too,gravel specimens experienced higher deviator stress thanthose of sand specimens. The triaxial test results show thatsand-gravel specimens with 30% sands experienced thehighest deviator stress showing the densest packing.

552

Gravel
50% sands

Deviator stress, q (kPa)

500

30% sands 15% sands

400
70% sands

300

Sand

200
100
0

Dr = 80%
0

2

4

6

8

10

12

0

15% sands
70% sands

Dr = 80%
0

2

4

6

8

10

12

14

Axial strain, εa (%)
Fig. 10: Deviator stress vs. axial strain from DEM
simulations with 80% of D r

Figs. 11 and 12 show comparisons of deviator stress vs.
axial strain between laboratory tests and DEM triaxial test
simulations for 50% and 70% of relative densities
respectively. Though Figs. 11 and 12 do not show a veryclear relation of different deviator stresses for the two
specimens (i.e., gravel and 30% sands specimens) for bothcases, Fig. 11 showed that the specimen with 30% sands
experienced higher deviator stress than the gravel specimen
in laboratory tests while Fig. 12 showed that the specimen
with 30% sands experienced higher deviator stress than thegravel specimen in DEM simulations.
500

Deviator stress, q (kPa)

30% sands,exp
400
Gravel,exp

300

Dr = 50%

200

0

1

2

3

4

5

6

7

0

1

2

3

4

5

Dr = 80%

6

7

8

9

10

Gravel

30% sands

48
44

Sand

GravelChange of φf with sands

30% sands

Change of φf with sands

40
36

Dr = 80%
Dr = 50%

Sand
1520

1600

1680

1760

1840

3

1920

2000

Dry density, ρd (kg/m )Fig. 13: Change of failure friction angle of sand-gravel
specimens with sands from laboratory experiments

Gravel,DEM
0Gravel,DEM

52

1440

30% sands,DEM
100

100

Fig. 13 shows failure friction angle, φ f , vs. dry density, ρ d ,of sand-gravel specimens from laboratory triaxial
compression tests. As shown in Fig. 13, the highest ρ d wasobserved for the specimen with 50% sands in both 50% and
80% of relative densities. In case of 80% of relative density,
the gravel specimen has the largest φ f , which gradually
reduced until the sand specimen. In case of 50% of relative
density, φ f increases until the specimen with 30% sands,
then reduced until the sand specimen. Fig. 14 shows failure
friction angle, φ f , vs. dry density for DEM simulated
specimens. As shown in Fig. 14, both dry density andfailure friction angle increased with sands (up to 30%
sands) and reached peak values for the specimens with 30%
sands in 80% of relative density. In case of 50% of relative
density, φ f remains same for the specimens with 0% - 30%
sands, then, reduced for the specimen with 50% sands andthen remains same until the sand specimen same as 80% of
relative case. As an overall finding, it is clear the specimen
with 30% sands showed densest behaviour, particularly, incase of 50% of D r in laboratory experiments and 80% of D r
in DEM simulations. The results showed that mixing ofsands in gravel (i.e., mixing of fines in ballast) change
failure frictional angle of the mixtures.
Failure friction angle, φf (degree)

Deviator stress, q (kPa)

Sand

40

30% sands,DEM

200

Fig. 12: Comparison of deviator stress vs. axial strain forthe specimens with 80% of D r

50% sands

80

300

Axial strain, εa (%)

Gravel

120

400

14

30% sands

160

30% sands,exp

500

0Axial strain, εa (%)
Fig. 9: Deviator stress vs. axial strain from laboratory
specimens with 80% of D r
200

Gravel,exp

600

Deviator stress, q (kPa)

600

8

9

10

Axial strain, εa (%)

Fig. 11: Comparison of deviator stress vs. axial strain forthe specimens with 50% of D r
553

Failure friction angle, φf (degree)34

5. ACKNOWLEDGMENT

Change of φf with sands

32

Gravel30

Japanese Government is highly acknowledged forproviding financial assistance to the first author to study in
Yokohama National University, Japan through a
Monbukagakusho scholarship.

30% sands

Sand

28
26

Change of φf with sands

Gravel
30% sands

24

Dr = 80%
Dr = 50%

22 Sand20

6. REFERENCES

1600

1680

1760

1840
3

1920

Dry density, ρd (kg/m )Fig. 14: Change of failure friction angle of sand-gravel
specimens with sands from DEM simulations

4. CONCLUSION
Triaxial compression tests were conducted on sand-gravel
mixtures to study effects of fine material mixings in ballast
in railway tracks. Triaxial tests were conducted in thelaboratory and using Yade, a DEM application as thenumerical simulations. Triaxial compression tests wereconducted on specimens having 50% and 80% of relative
densities. The following conclusions were made, Void ratios (both e max and e min ) reduced with initialaddition of sands. After reaching a minimum value,
void ratios increase with sands. Minimum value ofvoid ratios in DEM simulations were observed on thespecimens with 30% sands while it was 50% sands forlaboratory specimens.
 Void ratios observed in DEM simulations were smaller
than those of laboratory specimens. The difference invoid ratios can be attribute to difference in particle
shapes in DEM simulations and laboratory
experiments. However, relations of void ratio changewith sands were approximately same in DEM
simulations and laboratory specimens.
 The results also showed that void ratios can be simulated
in DEM simulations same as in laboratory experiments
for specimens with small particles (e.g., sands).
 As an overall finding, in both laboratory experiments
and DEM simulations, the highest deviator stress wasobserved on the specimens with 30% sands. The results
indicated that specimens with smaller void ratios
experienced higher deviator stress.
 However, there is difference in deviator stresses
between the two methods, experimental specimens
showing higher deviator stresses. The difference indeviator stress should be due to difference in particle
shapes in the two methods. However, the results
showed similar patterns in a qualitative manner.
 The specimens with 30% sands achieved the highest dry
density and maximum friction angle in DEM
simulations. In experimental specimens, it was quitesame for 50% of relative density.

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[13] Lin X and Ng TT, A three-dimensional discreteelement model using arrays of ellipsoids,
Geotechnique, vol. 47, No. 2, 1997, pp. 319-329.
[14] Yan G, Sui YH and Glenn M, Simulation of granularmaterial behaviour using DEM, In Proc., 6thInternational Conference on Mining Science &Technology, 2009, pp. 598-605.

554

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Rock slope stability problems in Gold Coast area, Australia
Shokouhi Ali, Gratchev Ivan, Charrismanagara Arry
Griffith University – Gold Coast campus - Australia


ABSTRACTThe Tamborine Mountain area of the Gold Coast, Australia often faces rock slope failures during rainy seasons. To havea better understanding of the factors that may affect the slope stability, several rock slope sites have been examined indetail to assess the effect of discontinuities formed in the rock masses, and degree of weathering of the rocks on theoverall stability of slopes. Different methods of slope stability analysis such as kinematic approach and slope mass
rating, were applied to investigate the potential failure mode and predict the behavior of rock slopes in the future. This
paper will present results of this work and make comparisons between different methods of rock slope stability analysis.
Keywords: slope stability, discontinuity, kinematic approach, SMR

1. INTRODUCTION
Tamborine Mountain, located in south of Brisbane andinland from the Gold Coast, is a plateau formed on
sub-horizontal basalt lavas of Tertiary age[1]. It is covered byforest. The basalt plateaux of southeast Queensland
potentially has hazard of landsliding[1], and often faces rockslope failures during rainy seasons.
Many of residences in Tamborine Mountain are located near
the rock slopes which have potential hazard of movement andinstability. Failure of these rock slopes can lead to damaging
results and will have a deep economic and social impact on
the society of this area. Also the slope failure can be very
harmful for infrastructures located in Tamborine Mountainsuch as roads. Every year during the rainy seasons, many ofaccess roads are closed by failed slopes.
The objective of this study is to investigate some cut slopes
through the main roads and analyze them with using
kinematic approach Slope Mass Rating (SMR) to identify thepotential failure modes of them and to determine thecondition of slopes located along roads.
The investigation field was located along Beaudesert –Nerang Road, immediately south-east of the Beaudesert –Nerang Road / Arunta Drive intersection, on the Gold Coast(Fig.1). The approximate length of the existing cut slope is
200m and with the height of approximately 7m. (Fig.2)

Figure 1 - Location on investigated slopes

Figure 2 - One of investigated slopes555

2. Geology and landslides types
The regional geological structure of the Tamborine Mountainand detailed geology of the basalt lavas forming the plateau
surface area have been described by Willmott[1], Cranfield etal [2], and Green [3].
The investigation field was located on the eastern part ofTamborine Mountain and made of metasediments of
Neranleigh-Fernvale Beds. Based on investigation performedby [1], this sequence of mainly sedimentary rocks was
deposited in Devonian to Carboniferous times. These rocks
have been deformed, indurated and slightly metamorphosed,
probably in the Carboniferous; they now can be termed
metasediments. The main rock types present are argillite(derived from siltstone and shale), greywacke dark sandstone
with abundant rock fragments), quartzite (derived from chert)
and greenstone (metamorphosed submarine basalt lavas).
Conglomerate horizons are known in a few areas. In someplaces, the argillite is sheared and has the appearance ofphyllite.
In 1981 a slope stability investigation performed by [1], on
Tamborine Mountain, his investigation showed that few
landslides are known on the plateau surface of the mountain,
but numerous slides, some very large, have occurred on thecliff lines and benches below, and on the fans of colluvial
debris along the lower western flanks of the mountain. Basedon his investigation, amongst different types of slope
instability, four main types of landslides can occur in
Tamborine mountain including: Rock falls, Debris flows,
small rotational slides and complex multiple rotational slides.

Set #

Dip

1
2
3

40
73
25

Location 3
ApertureDip(mm)direction
40
170
240

0.5
1-2
3-5

Compressive
Strength ofrocks (Mpa)
45
27
27

4. Data analyses
The collected data were analyzed by Dips computer software
to identify the potential failure modes of the investigatedslopes. The results from kinematic analysis indicate that at
location 1, there is only one potential failure mode. Planar
sliding is the most likely failure mode that may occur at thislocation (Fig.3). As can be seen in Fig3 some of the
discontinuities are located in gray shade that shows thepotential planar sliding area. Toppling and wedge failure
analyses show, there are no toppling and wedge potential
failure at location 1.

3. Field survey
To assess the impacts of formed discontinuity and weatheringon the rock slopes located in Tamborine Mountain, site
investigations were performed in 2012 to collect the relevant
and required data such as: Geometric properties of
discontinuities such as: dip, dip direction, spacing, etc.,
compressive strength by using Schmitt hammer and types ofrocks.
Three locations were selected along the rock slope located in
the side of main road for these investigations.
The following data were collected:
Figure 3 - Planar sliding of location 1

Table I- Collected data

Set #

Dip

1
2
3

55
50
55

Set #

Dip

1
2

61
73

Location 1
ApertureDip(mm)direction
220
0.5
135
0.5
50
2
Location 2
ApertureDip(mm)direction
70
170

1-2
1-2

Compressive
Strength ofrocks (Mpa)272727Compressive
Strength ofrocks (Mpa)
40
27

556

In location 2, the results produced by kinematic analysis
suggest that there may be a potential toppling failure (Fig.4).
In Fig.4, a part of the contour lines fall within the toppling
failure zone that indicates the potential toppling failure.

Figure 4 – Toppling failure of location 2

Figure 6 - Wedge failure of location 3

The results of kinematic analysis show that the rock slope at
location 3 has two potential failure modes, planar sliding andwedge sliding. In Fig. 5, the discontinuity set #3 falls within
the failure zone (grey shaded area) for planar sliding analysis.
This means that planar sliding may occur along this failure
plane.

Besides the kinematic analysis a SMR (Slope Mass Rating)analysis was also conducted to assess the slopes. Following
table indicates the results of SMR analyses.
Table II- SMR results
Location
1
2
3

SMR
40
83
14

Slope Mass Class
IV, Poor
I, Very good
V, Very poor

The results from SMR analyses consistently show thatlocation 2 has the best rock mass condition out of the threelocations, followed by location 1, and location 3 has theworst. These results reflect the evidence that there was asignificant recent rock fall at location 3. At location 1, therock fall was deemed to be minor, however, small fragmentsof rock blocks were found along the toe of the rock slope.

Figure 5 – Toppling failure for location 3
Fig.6 shows that there is a mean joint set orientationintersection (shown by black dots) that falls within the wedge
failure zone (grey shaded area). This indicates that a wedge
failure may potentially occur at location 3.

557

5. Concluding remarks
The results of kinematic analyses show that at location 1,
the rock is unstable under current conditions. However, in an
event of a rock slope failure, it will only be a minor one. At
location 2 of the rock slope, the geometry condition and highstrength of rock duo to less weathering lead to a better
condition than the other two locations. The SMR analysis
result of this location indicates that the rock slope is class ‘I’,
which means that the slope is completely stable. However, theresults from kinematic analysis suggest that there may be apossibility that the rock slope at this location will fail in
toppling.
All the analysis done on the rock slope at location 3, pointout that this section of the slope is the least stable out of thethree.
The following table illustrates the comparison of resultsbetween these two different approaches (Kinematic and
SMR).
Table III- Comparison of results
SMR
Kinematic approach
Set #
SMR
Grade
Toppling Planar Wedge
1
40
IV, Poor+2
83
I, Very good
+
3
14
+
+
V, Very poor
It can be deducted from the above table that, the results ofboth analyses have a good concordance with each other and interm of stability both show same outcomes.
6. REFERENCES
[1]
[2]
[3]

Willmott, W.F., Slope Stability and Its Constraints On Closer
Settlement On Tamborine Mountain, Southeast Queenslan, 1981,
Department Of Mines Geological Survey Of Queensland.
CRANFIE, L.C., SCHWARZBOCK, H., & DAY, R.W., Geologyof the Ipswich and Brisbane 1:250 000 Sheet areas, in Geological
Survey of Queensland, Report 95,1976.
GREEN, D.C., The volcanic rocks of Mt Tamborine, South-East
Queensland., in Department of Geology1964, University of
Queensland.

558

Techno-commercial development of pre-fabricated structures using
geopolymer concrete
J. Santhosh¹, R.Sanjeev², P.Sethu Raja³, A. Rajendra Prasadª¹
A. Abdul Rahmanª²,
Sri Sai Ram Engineering College, Chennai, India.
In most countries, concrete is widely used as thefoundation for the infrastructure. Concrete isused largely because its ability to support large
compressive loads. However, the use of concrete
is limited because it has low tensile strength. For
this reason, it is reinforced, and one of the more
popular reinforcing bars (rebar) is steel.

1. ABSTRACT:
It is known that cement production
releases high amount of CO 2 to the atmosphere.
The cement is used for civil construction
purposes including the structures like railway
sleepers, electric power poles, fencing and waste
containments, etc. In the present projectgeopolymer concrete is developed by using fly
ash, mixing with the alkaline material andhardening the structure by polymerization usingalkaline solutions like NaOH/KOH. A wastematerial such as rice husk ash which contains
silica is utilized.

Steel is available and is affordable in mostdeveloped countries but unfortunately not allparts of the world. In many countries, none orvery little steel reinforcement is used inconstruction. In some parts of the world manybuildings are constructed only with concrete or
mud-bricks. This is dangerous in case of seismic
activity. These buildings have little hope ofstanding in the case of an earthquake. Steelreinforcement would be an ideal solution, butcost is a considerable problem.

In the case of bamboo, a separate
treatment technology is developed to increase therigidity and flexibility in order to increase thelife span and eliminate its bio-degradability. Thisis done by using certain metal ions in the freshly
harvested bamboo. For making pre-fabricatedstructure, the bamboo is split into two halvesequally and geopolymer concrete is embedded
and the bamboo halves are united. The entire
pre-fabricated structure is cured over a period oftime.

3. MATERIALS AND METHODS:
The cement is a mixture of magnesium oxide(burnt magnesia) with magnesium chloride withchemicalformula Mg4Cl2(OH)6(H2O)8,corresponding to a weight ratio of 2.5–3.5 parts
MgO to one part MgCl2. It has the capacity towithstand about 10,000 to 12,000psi whereas theOrdinary Portland Cement has a capacity towithstand only around 2,000psi. MgCl2 is themost effective chemical which acts as dust
suppressant. The requirement of MgO and
MgCl2 are made available from the bittern, an
inexhaustible resource from the salt field and it iscost effective.

2. INTRODUCTION:
Many concrete structures have life span of two to
three decades. These structures deteriorate after
specific period. Many buildings are also
demolished after a period of 40 years. This
indicates that the raw materials used forconstruction last only for a specific period.
Under these circumstances, the cement
production increases.

Nearly 250 million tonnes of fly ash is being
dumped in developing and developed countries
each and every year. Developed countries
allocate high financial resource for dumping
these wastes. Since developing countries find it
difficult to allocate this high financial resource
this poses problems in the form of land use,
health hazards, and environmental dangers bothin disposal and in utilization. The other problems
are as contamination of ground water as it can

One ton of cement production releases equally
one ton of CO 2 thus contributing 7% of world
CO 2 emission that leads to global warming.
Cement is also known as most energy intensive
material.

559

leach into the subsoil, lowering of soil fertility,
siltation and clogging the system, reduction in
the pH balance, interference in the process ofphotosynthesis of aquatic plants and thus
disturbs the food chain, corroding exposed
metallic structures in its vicinity [1]. Fly ash
which has self-cementing properties is producedin bulk quantity from thermal power plants. It is
known to be useful for road construction and at
the same time, it is a good material that blendswith concrete [1].

4. BAMBOO
MATERIAL:AS

ARE-INFORCING

Bamboo reaches its full growth in just a few
months and reaches its maximum mechanicalstrength in just few years [3]. Its abundance intropical and subtropical regions makes it an
economically advantageous material. Some ofthe positive aspects such as a lightweight design,
better flexibility, and toughness due to its thinwalls with discretely distributed nodes and its
great strength make it a good reinforcing
material. Bamboo is used as structural material
for scaffolding at construction sites in India,
China and other countries as it is a tough,flexible, light weight and low cost material. In
nature when bamboo is covered with heavy
snow, it will bend until it touches the groundwithout breaking. This implies that bamboo has
greater flexibility than wood. The tensile strengthof bamboo is very high and can reach 380
N/mm2 [2]It contains about 50% of silica and around 14%
of ferric oxide which are some of the source
materials that imparts strength to concrete.
Ordinary Portland Cement contains only about
21% of Silica and 4% of Ferric Oxide. Fly ash
not only increases the strength but also increases
the volume since it acts as a dead weight material[2]. Class C fly ash which consists of about 20%
of CaO is used here. It has the property ofgaining strength over time.
The other source material used in the present
study is the ash of rice husk which contains highsilica content. It provides excellent thermal
insulation property which helps maintaining apleasant and cool environment.

It is estimated that one billion people live inbamboo houses. In India and China bamboo isused in construction of temporary suspension
bridges.

The process of using fly ash is to be evolved bymixing with alkaline material which is also
considered as waste material which will form agood product as cementious paste. This materialis hardened by adding chemical agents forpolymerization.
The
temperatureforpolymerization is around 60°C to 90°C. A
combination of NaOH or KOH with otherchemical components in the raw material getshardened like concrete. In addition, anotherbiological waste material such as rice husk isused.
In the process of polymerization, the alkaline
solution induces the Silica atoms especially inthe source materials like fly ash and/or husk ash.
In this process a gel is formed when it is appliedwith heat. This gel binds the aggregates of theunreacted
source material and this source
material due to polymerization at hightemperature forms a product known as
geopolymer concrete. This is used for mould
preparation.

4.1 BAMBOO CHARACTERISTICS:
Bamboo culms are a cylindrical shell divided bysolid transversal diaphragms at nodes and havesome intriguing properties such as high strength
in the direction parallel to the fibers, which run
longitudinally along the length of the culm, andlow strength in a direction perpendicular to thefibers. The density of fibers in cross-section of abamboo shell varies with thickness as well asheight. Fiber distribution is more uniform at thebase than at the top or the middle. This isbecause bamboo is subjected to maximum
bending stress due to wind at the top portion of
the culm. The strength of bamboo is greater thanmost of the timber products.
The mechanical properties vary with height andage of the bamboo culm. Research findings
indicate that the strength of bamboo increases
with age. The optimum strength value occurs
between 2.5 and 4 years. The strength decreases
at a later age. The function of the nodes is toprevent buckling and they play a role of axial
crack arresters.

560

One of the amazing aspects of bamboo is theway it interacts with the environment. It has been
discovered that bamboo can prevent pollution byabsorbing large amounts of nitrogen from waste
water and reducing the amount of carbon dioxidein the air.

Sodium dichromate, 300gms of Copper Sulphateand 150gms of Boric Acid in 10 litres of water.
The outer surface of the bamboo culms is hardand impermeous. Therefore the solution is notabsorbed. It is the inner surface of the culms
through which the solution is absorbed.

4.2 COMPARISION OF BAMBOO ANDSTEEL:

Bamboo culms can also be treated withPotassium Dichromate which makes it morerigid like steel rod. It can be used as a steel rod
and such structures are reinforced in geopolymer
concrete.

One of the properties that would make bamboo agood substitute to steel in reinforced concrete is
its strength. The strength of bamboo is greaterthan most timber products which areadvantageous. Bamboo is easily accessible as it
grows in almost every tropical and subtropicalregion; this lowers the cost of construction andincreases the strength of the buildings that would
otherwise be unreinforced.

The bamboo strips are also to be placed in a tankfilled with above solution. Soaking is done forabout 48 hours to make the treatment effective.
However, it is proposed to use electrically driven
compressor to impregnate the solution into thebamboo for future studies.the complete process
has been explained in Fig.1.

One major problem with bamboo is that it
attracts living organism such as fungi andinsects. Bamboo is more prone to insects thanother trees and grasses because it has a highcontent of nutrients. In order to combat thisproblem, it becomes necessary to treat bamboo
to protect it from bio-degradation. Steel does nothave this problem but it also needs to be coatedin order to protect it from rusting.
4.3 BAMBOO TREATMENT:The physical structure of bamboo makes it verystrong. Like reinforced cement concrete bamboo
represents a composite material. A a result ofstructural pecularities bamboo is very strong inspite of being light. Bamboo poles are treated toreduce their bio-degradability using 400gms of561

4.3.1 PROCESS PLANNING:

Fly AshRice Husk Ash

NaOH/KOH
For hardening

GeoPolymer Concrete

Mould
Preparation

Embedding Bamboo

CuringTesting the physical properties

Prefabricated Structure (Final Structure)

Fig.1

562

Treatment of Bamboo

excellent properties in both acid and baseenvironment. When curing period in oven is
prolonged and polymerization temperature isincreased, good structural property can beachieved. Life of structures made of geopolymer
concrete lasts for more than 6 decades.

5.1 PRODUCTION ADVANTAGES:
The Polymerization temperature is only about
90°C, hence furnaces with high refractoriness are
not required, thereby reduces the production costfurther. Geopolymer concrete is more resistant tocorrosion and fire, has high compressive andtensile strengths, and it gains its full strength
quickly (cures fully faster). It also shrinks lessthan standard concrete.

Concrete beams reinforced with unseasoned
bamboo showed slightly greater load capacities
than that with seasoned untreated bamboo. Highearly strength, low shrinkage, sulphate resistanceand corrosion resistance are some of thecharacteristics. Bamboo reduces the intensity ofdamage to buildings due to earthquake. Suchtreated bamboo is free from termite attack andacts as a biological steel rod.

5.2 SOCIAL IMPACT:
CO 2 emission reduces significantly thereby
reducing global warming. The local NGO and
Self Help Groups can be trained to fabricate theirown fencing poles.

6. DISCUSSION:

7. REFERENCES:[1] Malhotra, V.M., “Introduction: Sustainable
Development and Concrete Technology”,
Concrete International, V.24, No.7, July 2002,
p.22

Sorel Cement of Magnesium Oxide andMagnesium Chloride can withstand acompressive force of about 10,000 to 12,000psi
whereas the Ordinary Portland Cement canwithstand a pressure of only about 2,000psi. Fly
ash consists of about 50% of Silica and 14% of
Ferric Oxide which adds strength to the concrete
whereas it is only about 21% of Silica and 4% of
Ferric Oxide.

[2] Mehta, P.K., “Reducing the Environment
Impact of Concrete”, Concrete International,
V.23, No. 10, Oct, 2001, pp.61-66.
[3] Davidovitis, J. “Chemistry of Geopolymeric
Systems, Terminology”, Geopolymer’99
International Conference, France, 1999, pp. 9-40.

The load required to cause the failure of concretebeams reinforced with bamboo was from 4 to 5
times greater than that required for concrete
members having equal dimensions and with noreinforcement. Fabricated product is fire-proof
with low permeability, eco-friendly with

[4] Malhotra, V.M., “Making Concrete ‘Greener’with Fly ash”, Concrete International, V.21,
No.5, May 1999, pp. 61-66.

563

Mechanical Properties of Recycled Steel Fibre Reinforced Concrete
A.S.M. Abdul Awal1, Lim Lion Yee1, M. Dianah1 and M. Zakaria Hossain2
Faculty of Civil Engineering, Universiti Teknologi Malaysia, Johor Bahru, Malaysia
2
Graduate School of Bioresources, Mie University, Tsu-city, Mie, Japan

1

ABSTRACTThis paper presents some results, and discusses the feasibility of adding recycled steel fibre from scrap tires asreinforcement in concrete. A number of tests were conducted to investigate fresh and hardened state properties likeslump, Ve-be time, ultrasonic pulse velocity, compressive, tensile and flexural strength of recycled steel fibre
reinforced concrete (RSFRC). The effect of incorporation of recycled steel fibre on various aspect ratios (l/d) of 45,
67 and 89, and volume fraction (V f ) of 1, 1.5 and 2% were experimentally investigated. Test specimens comprisingof cube, cylinder and prism were prepared and tested after 1, 7 and 28 days of water curing. It has been found thatthe workability of RSFRC, in general, was reduced as the volume fraction of the fibre increased. The compressive
strength and splitting tensile strength of RSFRC reached a maximum at l/d ratio of 67. In terms of volume, thecompressive strength of RSFRC was not significantly increased by incorporation of recycled steel fibre. However,
the tensile and flexural strength of RSFRC was remarkably improved with the increase of fibre content. The results
obtained and the observation made in this study suggests that recycled steel from waste tires can successfully be
used as fibre reinforcing material in concrete.
Keywords: Fibre reinforced concrete, scrap tire, steel fibre, workability and strength.

1. INTRODUCTION
The technology of concrete reinforcement is not new, andfibres have been used for reinforcement since ancient times.
Fibre reinforced concrete is a type of concrete that includes
fibrous substances homogeneously dispersed and oriented in
the concrete matrix. Fibres commonly used in concrete aresteel fibers, synthetic fibres, glass fibres, and natural fibres.
Since the advent of fibre reinforcing of concrete, a great dealof testing has been conducted on various fibrous materials todetermine the actual characteristics and advantages for eachproduct. Over the decades different types of steel fibres havebeen successfully used to reinforce concrete.
Steel fibres may be produced either cutting wire, by shearing
sheets or from a hot-melt extract. The first generation ofsteel fibres was smooth so they did not develop sufficientbond with the cementitious matrix. Major efforts have been
made in recent years to optimize the shape and size of thesteel fibres to achieve improved fibre-matrix bondcharacteristics [1],[2]. It was found that steel fibre reinforced
concrete (SFRC) containing hook-ended stainless steel wires
has better physical properties than those containing straight
fibres. This is attributed to the better anchorage provide andhigher effective aspect ratio than that for the equivalent
length of straight fibre [3]. The addition of steel fibres toconcrete necessitate an alteration to the mix design tocompensate for the loss of workability due to the extra paste
required for coating the surface of the added steel fibres.
The utilization of SFRC over the past years has been sovaried and widespread. The most common applications arepavements, tunnel linings, bridge deck slab repairs, and soon. Unfortunately the fibres themselves are relatively
expensive. A 1% steel fibre addition, for instance, willapproximately double the material cost of the concrete, andthis has tended to limit the use of SFRC to special
applications. In recent years, there has been an increased
interest in using the recycled steel waste especially from tire
products. In all parts of the world large quantities of scrap
tires are generated each year resulting in environmental
hazards. It has been estimated that over 8 million units ofscrap tire have been generated in Malaysia, and 60% of thescrap tires are disposed via unknown route [4]. These scrap
tires are dangerous not only due to potential environmental
threat, but also from fire hazards and provide breeding
grounds for rats, mice, and mosquitoes. Today, scrap tire
disposal has become a serious issue and research works havebeen devoted to the use of steel fibres recovered from waste
tires in concrete. Concrete obtained by adding recycled steel
fibres has evidenced a satisfactory improvement of thefragile matrix, mostly in terms of toughness and postcracking behavior [5],[6]. This paper highlights the use ofrecycled steel fibre from scrap tires as reinforcement of
normal concrete and discusses various fresh and hardened
state properties of concrete containing the fibre.
2. MATERIALS AND TEST METHODS
The aim of this study as mentioned above, is to investigate
the properties of concrete reinforced with recycled steel
fibre recovered from waste tires. The fresh properties ofrecycled steel fibre reinforced concrete (RSFRC) wereslump and Vebe times while the hardened properties of werecompressive strength, splitting tensile strength, flexuralstrength and ultrasonic pulse velocity. The materials used
and the methods followed for conducting the tests are givenin the following sections.

564

2.1 Materials
Cement corresponding to ASTM Type I from a single source
was used throughout the experimental work. The coarse
aggregate was crushed granite with a maximum size of
20mm. Natural river sand with fineness modules of 2.5 wasused as a fine aggregate. Both coarse and fine aggregateswere batched in a saturated surface dry condition. Supplied
tap water was used throughout the study in mixing, curing
and other purposes.
A commercial polynaphthalene sulfonate type superplasticizer conforming to ASTM C494-92 was utilized ashigh range water reducing agent in the concrete. The dosage
of superplasticizer was kept constant for all concrete mixes
in order to eliminate any probable effect of this parameter on
the properties of hardened concrete.
2.2 Preparation of Recycled Steel Fibre
Steel fibres used in this research were recovered by ashredding process of waste tires. The steel was successively
separated from rubber by a pyrolysis process. The steel
fibres shown in Fig.1 are characterized by different
diameters and length. The average fibre length was 30mmwith an average tensile strength of 1033 MPa.
A preliminary characterization was carried out by astatistical analysis in order to evaluate the variability of thegeometric properties after the shredding process without any
further treatment. The characterization is fundamental toeventually defining the opportune treatments for improving
the final properties of the concrete. A sample of about 200
steel fibres, extracted randomly after the shredding process,was analysed. The diameter of each fibre was recorded by amicrometer and determined averaging three measures,
namely at the two extremities of the fibre and at the
midpoint.
2.3 Concrete Mix and Sample Preparation
The recycled steel fibres were added at different aspect ratio(l/d) and volume fraction (V f ) to determine the applicabilityof the fibre as reinforcement in concrete. The steel fibre
recovered from waste tyre was employed to concrete
mixtures with aspect ratios (l/d) of 45, 67 and 89 followedby volume fractions of 0.5%, 1.0%, 1.5% and 2.0%. A total

Before shredding

After shredding

Fig. 1: Recycled steel fibre before and after shredding.
of 63 cubic specimens, 21 cylindrical specimens and21 prismatic specimens was fabricated and cured in water
until testing. Water to cement ratio was settled to 0.4
according to ACI Committee such as to provide goodmechanical strength and adequate workability of the
mixtures. The mix proportion of recycled steel fibre
reinforced concrete is shown in Table 1.
The mixing procedure was divided into three stages. In thefirst stage cement, fine and coarse aggregates were weighed
and mixed by hand until all the constituents are mixed
uniformly. In the second stage, the recycled steel fibres wereadded slowly and uniformly to eliminate the recycled steel
fibres clumped together. In the final stage, measured waterand superplasticizer were added and mixed thoroughly untila homogeneous mix is obtained.
2.4 Tests on Fresh and Hardened Concrete
Slump test, following ASTM C143 standard was carried out
to measure the consistency of a concrete which has a closeindication to workability. Along with slump test, the Vebe
test was conducted conforming to ASTM C 1170. In order toinvestigate the compressive strength of concrete, uniaxial
compression test was carried out on 100 mm cube specimen
following BS 1881-127:1990. The splitting tensile test was,however, performed on the standard test cylinders
measuring 100x200mm conforming ASTM C496/C496M04. The flexural strength test was conducted using
100x100x500mm beams under third point loading following

Table 1: Mix design of recycled steel fiber reinforced concrete.
Mixturecode

l/d ratio

V r (%)

Cement(kg/m3)

Water(L/m3)

Coarseaggregate(kg/m3)Fine
aggregate(kg/m3)Fibre(kg/m3)

SP(kg/m3)

PC

-

-

562

225

925

670

-

5.63

RSFRC -1

45

0.5

562

225

925

670

12.5

5.63

RSFRC -2
RSFRC -3

67
89

0.5
0.5

562
562

225
225

925
925

670
670

12.5
12.5

5.63
5.63

RSFRC -4
RSFRC -5

67
67

1.0
1.5

562
562

225
225

925
925

670
670

25.0
50.0

5.63
5.63

RSFRC -6

67

2.0

562

225

925

670

70.0

5.63

565

the ATSM C1609/C1609M -10.
In addition to compressive, tensile and flexural strength
tests, an ultrasonic pulse velocity test (conforming ASTM
C597) was also conducted to measure the uniformity ofconcrete, and to investigate the relationship between
compressive strength and ultrasonic pulse velocity ofconcrete cubes with different fibre volume fractions.
3. RESULT AND DISCUSSION
3.1 Properties of Fresh Concrete
Both the slump and Vebe time tests were conducted toinvestigate workability of concrete. The slump value and theVebe time of all the concrete mixes for different aspect ratio(l/d) and fibre volume fraction (Vf) are tabulated in Table 2.
As shown in the Table 2, slump values of concrete decreased
as the l/d ratio and Vf increased. Meanwhile, Vebe time of
concrete increased as the l/d ratio and Vf increased. For fibre
volume of 1% and above the workability of concrete is
dramatically decreased. The interlocking of fibres resists theflow of fresh concrete affecting the workability of concrete.
The result on unit weight of concrete, presented in the sametable reveals that the unit weight of concrete increased
uniformly with the increase in fibre volume fraction. This is
obvious as the specific gravity of recycled steel fibres arehigher than those of other components of concrete.
Regardless of fibre volume, it was further observed that theunit weight of RSFRC decreased by the increase of fibre
aspect ratio because of air content in the concrete byorientation and distribution of long fibres.
Aspect ratio (l/d) of recycled steel fiber has a predominant
effect on the workability concrete mixtures. The slump value
has been found to decrease in concrete mixture realized withfiber aspect ratio of 45 and 67. A significant drop from50mm to 30mm for concrete mixture was realized with fiber
aspect ratio of 89. Table 2 reveals that higher the aspect ratio
higher was the Vebe time for the same fibre content. This isdue to the fact that long fibres tend to mat together while
short fibres does not interlock and can be dispersed easily byvibration.

The experimental results obtained in this study are in close
agreement with the research findings from Yazici et al. [8].
It has been shown that slump of concrete decreased as the l/dratio of steel fibre increased. Like that of slump value, theVebe time also found to be influenced by the aspect ratio ofrecycled steel fibers.
Like aspect ratio, the amount of fibre i.e. the volume fration(V f ) significantly influenced the workability of fibre
reinforced concrete. Higher the amount of fibre lower was
the slump value. With an increase in the amount of fibre for
1 to 2 percent, the Vebe time of the SFRC has been found tobe almost double. This is to be expected because lower
amount of fibre can easily disperse in the concrete matrix.
With further addition the fibres start to clump together
showing a balling effect (Fig. 2) Furthermore, theinterlocking and entangled around aggregate particles andconsiderably reduce the workability. A similar observation
has been made by Uygunoglu [7] who found that the unit
weight of SFRC increased with the increase in fibre content
for both 30mm and 60mm steel fibre.

Fibre Concentration

Fibre Balling

Fibre Concentration

Fibre BallingFig. 2: Fresh concrete showing fibre concentration andballing effect.
3.2 Hardened Properties of Concrete
The mechanical properties of recycled steel fibre reinforced
concrete investigated in this study are compressive strength,
splitting tensile strength and flexural strength. The results

Table 2: Slump, Vebe time and unit weight of concrete mixtures.
Mixturecode

l/d ratio

Vf (%)

Slump(mm)

Vebe time
(s)

PC

-

-

50

3.05

2

RSFRC-1

45

0.5

45

3.88

2

RSFRC-2

67

0.5

40

4.06

2

RSFRC-3

89

0.5

30

5.62

2

RSFRC-4

67

1

25

6.35

2

RSFRC-5

67

1.5

20

7.85

2

RSFRC-6

67

2

10

12.32

2

566

Unit weight3(kg/m )

3.2.1 Compressive Strength
The 28-days compressive strength values are 57.15, 58.50
and 57.85MPa respectively. Among the three aspect ratios,
the l/d ratio of 67 (RSFRC-2) exhibited the maximum
compressive strength value. It can be seen that on average,
compressive strength of RSFRC are about 2-6% higher thanthat of control mixture. RSFRC with fibre volume of 1.5%
showed the highest compressive strength for aspect ratio of
67. A close observation has been made by Modtrifi et al. [9]who reported that an addition of 0.7% recycled steel fibres
in concrete resulted in an increase of approximately 12% incompressive strength.
Although the compressive strength of SFRC was muchaffected by the presence of fibre, the failure mode, however,exhibited a considerable change from fragile to ductile state.
Due to bridging effect of the fibre, the cubic specimens did
not crush but held their integrity up to the end of the test.
Fig. 3 illustrates the typical failure mode of plain concrete
and recycled steel fibre reinforced concrete.

There appears to be a good correlation between compressive
strength and ultrasonic pulse velocity containing recycledsteel fibre. The relationship shown in Fig. 4 suggests thathigher the compressive strength higher was the ultrasonicpulse velocity. This is in agreement with the researchfindings in concrete containing recycled aluminium can fibre[10] and the relationship demonstrated by Neville andBrooks [11].
70Compressive Strength (MPa)obtained for all categories of strength investigation arepresented in Table 3. Along with strength, ultrasonic pulse
velocity of the concrete specimen was also studied to
examine the relationship.

60
50
40
30

0%
0.50%

20

1.00%
1.50%

10
0
2100

2.00%

2200
2300
2400
2500
Ultrasonic Pulse Velocity (m/s)

2600

Fig. 4: Relationship between compressive strength andultrasonic pulse velocity of concrete with different amountof fibre.
3.2.2 Splitting Tensile Strength

Normal concrete

RSFR concreteFig. 3: Typical failure mode of normal and recycled steel
fibre reinforced (RSFR) concrete cube.

In general, the tensile strength of RSFRS was found toincrease with increasing amount of fibre. But in all mixes,
concrete with fibres volume fraction of 0.5%, l/d ratio of 67(RSFRC-2) showed the maximum strength gain. The
relative splitting tensile strength values of RSFRC according
to V f and l/d ratio are also given in Table 3. The test results
indicated that splitting tensile strength of RSFRCs is about

Table 3: Mechanical properties of concrete mixtures.
l/d ratio

Vf(%)

Compressive
strength, fc(MPa)

Relative , f c(%)

Splittingtensilestrength, fst(MPa)

Relative, f st(%)

FlexuralStrength, f f(MPa)

Relative,
ff
(%)

PC

-

-

55.85

100

3.80

100

4.60

100

RSFRC-1

45

0.5

57.15

102

6.50

172

5.20

113

RSFRC-2

67

0.5

58.50

105

6.90

181

5.45

118

RSFRC-3

89

0.5

57.85

104

6.65

175

5.65

123

RSFRC-4

67

1

58.20

104

7.10

187

5.75

125

RSFRC-5

67

1.5

59.00

106

8.15

215

6.15

134

RSFRC-6

67

2

58.70

105

9.45

249

6.95

151

Mixturecode567

70-149% higher than the control mixture. Indeed, recycledsteel fibres significantly improved the splitting tensile
strength of concrete as compared to compressive strength.
The results obtained in this study are consistent withprevious studies [9], [10], [12].
During the splitting tensile test, the effect of the recycledsteel fibres was apparent. The recycled steel fibres appear tocontrol the cracking of RSFRC and alter the post cracking
behaviour. The recycled steel fibres seem to provide a load
redistribution mechanism after initial cracking. Unlike innormal concrete, it was difficult to separate the fractured
specimens because the recycled steel fibres were bridging
the gap that kept the two concrete parts together, as shownin Fig. 5.

Fig. 6: Failure mode of plain concrete beam in flexure.

Fig. 7: Failure mode of RSFRC beam in flexure.
4 CONCLUSION

Fig. 5: Concrete cylinder after splitting tensile test.

The experimental works carried out in this research draw
some conclusions. These are:3.2.3 Flexural Strength
Data shown in Table 3 reveals that flexural strength of
RSFRC are about 14-52% higher than the control sample.
The flexural strength of control specimen, for example, is4.60MPa. For l/d ratio of 45 (RSFRC-1), 67 (RSFRC-2) and89 (RSFRC-3) the flexural strength values are 5.20, 5.65 and5.45MPa respectively. The relative flexural strength values
of RSFRC according to V f and l/d ratio also given in Table
3. Evidently the flexural strength of concrete hassignificantly been improved by incorporating recycled steel
fibres. Higher the aspect ratio and volume fraction higher
was the strength gain.
An important aspect that can be outlined from experimental
results is the ductility and tenacity of RSFRC increase whenfibre content volume is increased, and for the same fibre
content, when fibre length is increased. This phenomenon
could be due to the higher deformability and energy
absorption of RSFRC during the cracking phase.
Interestingly the RSFRC shows a higher bending stiffness
and a different cracking pattern than the normal concrete.
Most cracks showed a reduced width and they appeared
from casting surface due to the presence of recycled steel
fibres, which tend to move to the lower portion of thespecimen. The differences in the failure mode of plain and
RSFRC are illustrated in Fig. 6 and Fig. 7.

1) The inclusion of recycled steel fibre significantly
affected the workability of concrete. It has been foundthat the workability of concrete mixes decreased with
the increase in the aspect ratio (l/d) and the volume
fraction (Vf) of fibre.
2) Recycled steel fibre has been found to improve theoverall strength of concrete. This improvement has beenshown to be prominent in case of splitting tensile andflexural strength as compared to that of compressive
strength. Generally, higher the l/d and V/f, higher was
the strength development.
3) There has been a good co-relation between compressive
strength and ultrasonic pulse velocity of concrete. The
ultrasonic pulse velocity of concrete was found toincrease with the increase of compressive strength.
4) The result obtained and the observations made in thisstudy suggest that the recycled steel fibre from waste
tire has a good potential as reinforcing material. Longterm research work including deformation behavior anddurability aspects of concrete, however, has been put
forward for future study to obtain better understandingof this material in concrete.

568

REFERENCE[1] Mindess, S, Young, FJ, Darwin, D. Concrete. New
Jersey. Pearson Education, 1996.
[2] Colin, DJ. “Steel fibre-reinforced concrete – present andfuture in engineering construction”. Composites, 1982,
pp. 113-121.
[3] Swamy, RN, Mangat, PS, “The interfacial bond stress insteel fiber cement composites”. Cement and Concrete
Research, vol. 6, 1976 pp. 641-650.
[4] Jalil, MA, “Sustainable development in Malaysia: A
case study on household waste management,” J. of
Sustainable Development, vol. 3, No. 3, 2010, pp. 91102.
[5] Tlemat, H, Pilakoutas, K, Neocleous, K, “Design issues
for concrete reinforced with steel fibre recovered fromwaste tires,” J. of Materials in Civil Engineering, ASCE
18 (September/October), 2006, pp. 677–685.
[6] Naik, TR, Singh, SS. Utilization of discarded tires asconstruction materials for transportation facilities.
Report No. CBU-1991-02, UMW Center for byproducts utilization, University of WisconsinMilwaukee.1991, pp.16.
[7] Uygunoglu T, “Effect of fibre type content on bleedingof steel fibre reinforced concrete,” Construction and
Building Materials, 2010, pp. 766-772[8] Yazici S, Inan G, Tabak V, ”Effect of aspect ratio andvolume fraction of steel fibre on the mechanicalproperties of SFRC,” Construction and Building
Materials, vol. 21, 2007, pp. 1250-1253.
[9] Modtrifi N, Ibrahim IS, Rahman AB, “Mechanicalproperties of concrete containing recycled steel fibres,”in Proc. 3rd Int. Conf. of EACEF (European Asian Civil
Engineering Forum) Universitas Atma Jaya Yogyakarta,
Indonesia, September 20-22, 2011, pp. B-135 - B140.
[10] Awal, ASMA, Mazlan, D, Mansur, ML, “Ultisation ofsoft drink can as reinforcement in concrete” in Proc. 3rd
Int. Conf. of EACEF (European Asian Civil
Engineering Forum) Universitas Atma Jaya Yogyakarta,
Indonesia, September 20-22, 2011, pp. B15-B20.
[11] Neville, AM, and Brooks, JJ, Concrete Technology,
ELBS Edition, Longman Group UK Ltd., 1994.
[12] Papakonstantinou, CG, and Tobolski, MJ, “Use ofwaste tire steel beads in Portland cement concrete,”
Cement and Concrete Research, vol. 36,
2006,
pp.1686-1691.

569

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effects of Gradation of Various Sand Deposits of Pakistan on Strength ofHardened Concrete
Gauhar Sabih 1, S. Muhammad Jamil1 and Kamran Akhtar1
School of Civil and Environmental Engineering, National University of Science and Technology, Islamabad,
Pakistan1

ABSTRACT
Various sand deposits of varying gradation are available in different regions of Pakistan. The gradation of different
sands was compared with the ASTM specifications. It was revealed that only two sands satisfy the ASTM
gradation limits. The same was also confirmed by the results of workability and compressive strength. For
betterment of the construction industry, optimization of sand was carried out. The grading curves of optimized
sands became compatible with the ASTM specifications and the concrete made with improved sands showed up to
39% increase in compressive strength. The optimized sands were cost effective also.
Keywords: sand, concrete, Pakistan, Lawrencepur, stone dust, compressive strength
1. INTRODUCTION
Since up to 30 percent of the total volume of concreteconsists of sand/fine aggregate consequently fineaggregate characteristics significantly affect theperformance of fresh and hardened concrete and have an
impact on the cost effectiveness of concrete [1].
Gradation or particle size distribution of sand affects theproperties of concrete like packing density, voids content,
workability and strength [2]. Uniformly distributed
mixtures generally lead to higher packing resulting in
concrete with higher density and less permeability, andimproved abrasion resistance. Consequently, uniformly
distributed mixtures require less paste, thus decreasing
bleeding, creep, and shrinkage [3]-[4]. The fine aggregate
should be uniformly graded. If fine aggregate is toocoarse it will produce bleeding, segregation andharshness, but if it is too fine, the demand for water will
be increased [5].
Various sand deposits of varying gradation are available
in different regions of Pakistan. The existence of large
water reservoirs usually transports finer sands towards
downstream; consequently the sand in the Northern
region is coarser as compared to the sand from Central
and Southern regions. There is a general trend in the
country to use coarser sand from Lawrencepur, a site
located in the Northern region. Transportation cost ofsand from this place to far flung areas in Centre and Southmakes any project cost intensive. This study was aimed atevaluating the grain size distribution and absorption ofvarious types of sands in the light of ASTM standards andto contrast the effects of gradation on the strength andworkability of concrete. Finally devising a technique for
optimizing the gradation of fine sands by sieving thestone dust and mixing different fractions of stone dustwith the natural sand to get optimized sand gradation andevaluating its effects on strength of hardened concrete.
2. IDENTIFICATION OF SAND DEPOSITS
Various well known material suppliers and construction
firms were approached for identification of major sanddeposits of the country. The locations of identified sanddeposits are shown in Fig 1.
3. EXPERIMENTAL PROGRAM
3.1 Sieve Analysis and Classification
All the sand samples were classified by sieve analysis,
carried out according to ASTM C-136-96 [6] and theclassification of sand was done according to ASTM D2487-93 [7]. ASTM C-33-03 specifies the gradation limits
for sands to be used as fine aggregate in concreting [8].
The gradation curves of all the sands were compared with
the ASTM gradation limits to contrast the variations.
3.2 Absorption Percentage
All the sand samples were tested for their absorption
percentage in saturated surface dry (SSD) condition
according to ASTM C-128.
3.3 Organic Impurities in Sands
This test was conducted for all the sands according to
ASTM C-40 [9]. The test procedure consisted of making areference standard colour solution by dissolving reagent
grade potassium dichromate in concentrated sulfuric acidat the rate of 0.25 gram per 100 ml of acid. The sandsample was filled in a glass bottle up to 130 ml mark andsodium hydro oxide solution (3% by mass in 97% water)

Fig 1. Locations of Identified Sand Deposits
was added up to 200 ml mark. The bottle was shaken
vigorously and allowed to stand for 24 hours. After 24hours the colour of the sand solution was comparedwith freshly made reference colour solution. The darker
colour than reference colour indicates the possibility ofsand containing injurious organic impurities andrequires further testing for sand before use inconcreting. The lighter or equal colour means that thesand does not contain injurious organic impurities andno further testing is required.
3.4 Evaluation of Sands in Concreting
3.4.1 Concrete Mix
The concrete mixes were prepared using sand / fineaggregate from eight different sources and remaining
variables as constant with the cement: sand: coarse
aggregate ratio of 1: 2: 4 (by weight) respectively,
cement content (201.7 kg/cum) and water cement ratioof 0.5 with constant slump value. The course and fineaggregates were used in SSD condition. The course
aggregate used was taken from Margallah quarry
located in the Northern region. The cement used was
Fauji cement (type 1, portland cement). The water used
was normal tap water without any undesired colour.
3.4.2

Workability of Fresh Concrete

The workability was evaluated by measuring the slumpof fresh concrete according to ASTM C-143. The
concrete was filled in the slump cone in 3 equal layerswith 25 blows of tamping rod each. After leveling the

top layer, the cone was lifted and slump was measured
as reduction in original height.
3.4.3 Compressive Strength of Hardened Concrete
The test was conducted on 4x4 inches cubes (as per BS1881) and 4x8 inches cylinders (as per ASTM C-39).
The compressive strength was determined at the age of
3, 7, 14 and 28 days [10]-[11].
4.

TEST RESULTS AND DISCUSSIONS

4.1 Sieve Analysis and Classification
All the sand types are poorly graded sands classified as
SP, except the Wadd sand which contains silt as welland classified as SP-SM. The Percentage of micro fines(material passing No. 200 sieve) is below 5% in all
sands except Wadd sand which contains 9% microfines. According to ASTM C-33, the allowable limit of
micro fines in sand for concreting is up to 5% so all the
sands satisfy the ASTM criteria except the Wadd sand.
4.2
Comparison
Gradation Limitsof Gradation

with

ASTM

The comparative results of all the sands with the ASTM
gradation limits are shown graphically in Fig 2. All the
sands are finer than ASTM gradation limits except the
Qiblabandi and Bollari sand. The Qiblabandi sand is thebest match to ASTM gradation limits while Bollari sandis little finer than Qiblabandi sand.

5. OPTIMIZATION OF SAND GRADATION

Fig 2. Sand Gradation versus ASTM Gradation
Limits
4.3 Absorption Percentage
All the sands have absorption less than 3 % so they arewithin range of the limiting range as per BS-8007. The
Qiblabandi sand has the lowest absorption percentage
i.e. 1.75% being the coarsest sand of all types while the
Bollari sand has the highest absorption percentage of
2.11% but it is also within range of the acceptable limit.

Having observed the gradation of various sands, anendeavor was made to bring them within the specified
limits of ASTM by adding a fraction of coarse material.
The stone dust from Margallah, crusher site was taken
and various fractions of stone dust were produced bysieving the stone dust on Number 4, 8, 16, 30 and 50
sieves (stone dust can be taken from any crusher site as
the lab tests showed compatible results on use of stone
dust from various crusher sites). Different percentagesof various fractions of stone dust were added to the sandsamples and optimized samples were produced. The
optimized samples were than tested for their gradation
to check the compatibility with ASTM limits. The
optimized samples were then used for making concrete
cubes and cylinders thus determining the improvementin the compressive strength of concrete specimens.

4.4 Organic Impurities in Sands
All the sand sample solutions have the colour lighter
than the reference standard colour solution so all the
sands do not contain injurious organic impurities and nofurther testing is required for the organic impurities(according to ASTM C-40). Consequently all the sands
are suitable for concreting as per the results of this test.
4.5 Workability of Fresh Concrete
The slump values of the concrete made with different
types of sands are ranging between 1.1 to 1.2 incheswith same water cement ratio. This slump value wastaken as constant for further compressive strength
testing of hardened concrete. According to
specifications, hand tamping was resorted in the
preparation of cubes while vibrator insertions were usedin preparation of cylinders [12]-[13].
4.6 Compressive Strength of Concrete Cylinders
and Cubes
The compressive strengths of both of concrete cubes
and cylinders follow the same trend line thus
confirming the results. The comparative results ofcompressive strength of concrete cylinders aregraphically shown in Fig 3. The Qiblabandi sandproved to be the best fine aggregate giving highest
compressive strength of 3463 psi. Gradation meeting
the ASTM specifications, lowest percentage of microfines and within range fineness modulus of Qiblabandi
sand are the supporting factors for the highest
compressive strength. The trend of gradation ofdifferent sands as compared with the ASTM gradation
limits is also confirmed by the results of thecompressive strength. The Ghazi, Ravi and Chanab
sands are below 3000 psi which is not desirable forconcreting works.

Fig 3. Comparison of Compressive Strength ofConcrete Cylinders (psi)This resulted in significant improvement of sandgradation as well as the compressive strength of
concrete. Six sands which were out of range of ASTM
standards were optimized.
Various different
combinations were tried for making the improved sandsamples to attain the best mix.
5.1 Gradation of Optimized Sands
The gradation curves of optimized sands werecompared with the gradation curves of original sands inthe light of ASTM gradation limits to determine theeffects on the gradation curves. The comparison clearlyshows that the gradation curves of optimized sands are
satisfying the ASTM gradation limits where as theoriginal ones were out of the limits. The comparison oforiginal and optimized gradation curves is shown in Fig
4.
5.2 Compressive Strength of Concrete
The optimized sand samples were tested by making
concrete cubes and cylinders with the cement: sand:coarse aggregate mix of 1: 2: 4 (by weight) and thevariables like cement (201 kg/cum), water-cement ratio(0.5), curing conditions and course aggregate(Margallah) as constant. The concrete specimens werethan tested for compressive strength at the age of 3, 7,
14 and 28 days to get the improved compressive

strengths. These improved strengths were thancompared with the original compressive strength values
to get the percentage improvement.

6.1 Comparative Results of Cylinder Strength
The comparative summary is shown graphically in Fig
6. The comparison of percentage improvement of
optimized sands in comparison to original sands isshown in Fig 7. The results show that all the optimized
sands gave increased compressive strength. The
optimized Ravi and Chanab sands show the mostsignificant increase of 37% to 39% in the compressive
strength when compared with original sands.
7. CONCLUDING REMARKS
Based on laboratory investigations and data analyses,
following conclusions are drawn:•

Fig 4. Comparison of Optimized Gradation Curves
with ASTM Limits
5.3 Results of Compressive Strength of Optimized
Sands
The compressive strengths of both cubes and cylinders
follow the same trend line thus confirming the results.
The comparative results of compressive strength of
concrete cylinders are graphically shown in Fig 5. The
results show that there is significant improvement in thecompressive strengths and all the improved sands areabove 3000 psi which is the foremost requirement for asand to be used in concreting works. The improved
Lawrencepur sand and Durrani sand (with finenessmodulus of 2.62 and 2.58) respectively are giving thehighest strength of 3441 psi and 3547 psi respectively.
6.

COMPARISONOFORIGINALAND
OPTIMIZEDSANDPROPERTIESBASED ON COMPRESSIVE STRENGTH
OF CONCRETE

The cylindrical compressive strengths of optimized
sands at the age of 28 days were than compared withthe cylindrical compressive strength values of original
sands to get the percentage improvement.

•
•

•

•

•

Local sands can be utilized in the construction
projects by optimizing the gradation usingstone dust to avoid high costs of transportation.
Even the best considered sand deposit requires
fractions of stone dust to maximize its
potential in concreting.
The optimized Ravi and Chanab sands show
the most significant improvement of 37% to
39% in the compressive strength while theoptimized Lawrencepur, Ghazi, Wadd andDurrani sands also show greater strength thanoriginal sands with percentage improvement of
8% to 16%.
The optimized Ghazi, Ravi and Chanab sands
can be utilized in concreting works where as
they were not suitable for concreting in their
original form.
All the optimized sands show higher
compressive strength than the original sands inconcrete cubes and cylinders both and areabove 3000 psi which is the minimum criteria
for any sand to be used as fine aggregate inconcreting works.
The improved performance of optimized sands
can be directly related to various factors out of
which, improved sand gradation is a major
factor as per the analysis of this study.

Fig 6. Comparison of Original versus Improved
Cylinder Strength
Fig 5. Comparison of Improved Compressive
strength of Cylinders (psi)for Testing and Materials, 2000.
[11] British Standards Institution, "Methods for makingof concrete cubes in laboratory," BS-1881-116: Part
110, 1990.
[12] British Standards Institution, "Methods for
Determination of compressive strength of concrete
cubes," BS 1881-108: Part 110, 1990.
[13] ASTM C-39, Standard Specification for
Compressive Strength of Cylindrical Concrete
Specimens, Philadelphia, PA: American Society for
Testing and Materials, 2003.

Fig 7. Comparison of Percentage Improvement of
Optimized Sands
ACKNOWLEDGEMENTS
The authors thank Dr. S. Ali Rizwan for his criticalreview which greatly helped in improving themanuscript. The study was funded by the National
University of Sciences and Technology.

REFERENCES[1] Hudson, B.P, “Concrete Workability with HighFines Content Sands”, 1999.
[2] Golterman, P., Johansen, V., Palbfl, L., “Packing ofAggregates: An Alternative Tool to Determine theOptimal Aggregate Mix,” ACI Materials Journal,
Vol. 94, No. 5, 1997.
[3] Shilstone, J. M., “The Aggregate: The
Most
Important Value-Adding Component in Concrete,”
Proceedings, Seventh Annual International Center
for Aggregates Research Symposium, Austin,
Texas, 1999.
[4] Washa, G.W., Workability, Chapter 5,
Concrete
Construction Handbook, 1998, 4th edition, NewYork.
[5] Galloway, J. E. Jr., “Grading, Shape, and Surface
Properties,” ASTM Special Technical Publication
No. 169C, Philadelphia, 1994.
[6] ASTM C-136, Standard Test Method for
Sieve
Analysis of Fine and Coarse Aggregates,
Philadelphia, PA: American Society for Testingand Materials, 2001.
[7] ASTM D-2487, Standard classification of soils forengineering purposes (Unified soil classification
system), Philadelphia, PA: American Society for
Testing and Materials, 1993.
[8] ASTM C-33, Standard Specification for Concrete
Aggregates, Philadelphia, PA: American Society for
Testing and Materials, 2003.
[9] ASTM C-40, Standard test method for organic
impurities in sand for concrete, Philadelphia, PA:
American Society for Testing and Materials, 1992.
[10] ASTM C-192, Standard Practice for Making andCuring Concrete Tests Specimens in the
Laboratory, Philadelphia, PA: American Society

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Assessment of recycled concrete aggregates for road base and sub-base
S. Jayakody, C. Gallage, A.Kumar
School of Urban Development, Faculty of Built Environment & Engineering, Queensland University ofTechnology, Brisbane, Australia
ABSTRACT
Population increase and economic developments can lead to construction as well as demolition of infrastructures suchas buildings, bridges, roads, etc and used concrete is the main waste product of them. Recycling of waste concrete toobtain the recycled concrete aggregates (RCA) for base and/or sub-base materials in road construction is a foremost
application to be promoted to gain economical and sustainable benefits. As the mortar, bricks, glass and asphalt present
in different constituents in RCA, it exhibits inconsistent properties and performance. In this study, six different types of
RCA samples were subjected classification tests such as particle size distribution, plasticity, compaction test andCalifornia Bearing Ratio (CBR). Results were compared with those of the standard road materials used in Queensland,
Australia and found that ‘RM1-100/RM3-0’ and ‘RM1-80/RM3-20’ samples are sitting in the margin of the minimum
required specifications of base materials while others are lower than that.
Keywords: Recycling, Waste concrete, Road materials
1. INTRODUCTION
Demolished materials are becoming more popular to recycle
and reuse due to shortage of natural mineral resources,
increasing waste disposal cost, and increasing the demand of
materials. Conversely supplying the conventional aggregatesfor construction purposes make impact on resource depletion,
environmental degradation, and energy consumption.
Therefore reusing the recycled materials creates many
economical and environmental benefits.
Crushed concrete can be considered and promoted as analternative and a sustainable source of aggregate forconstruction industry. Successful studies have been revealed
that recycled concrete aggregates can be applied as a partial
or complete substitution of natural aggregates in the
production of ordinary concrete [1], [2], [3]. However Hansen
in 1992 [4] has proven that fine portion of RCA makes
detrimental effects on the harden properties of concrete andthe coarse RCA are the best use.
The studies on RCA as pavement materials are widelyreported in last decades. Reference [5] shows the evaluationof the performance of RCA using as a base material under hot
mix asphalts and as an aggregate in Portland cement concrete
pavement while [6] reported that the performance of thesubbase materials prepared with both course and fine RCA
was comparable to that of conventional subbsase materials.
Reference [7] introduced RCA as base and subbase material
for supporting a concrete pavement while [8] concluded thefeasibility of using RCA and crushed clay bricks as blended
materials for un-bound granular subbases. Although theseresearch findings have demonstrated the feasibility of using
RCA assubbase material as well as for base course of the concrete
pavements, a detailed investigation of RCA as base andsubbase material for unbound pavements is required
including classifications considering the variability of RCA
compositions, performance characteristics under repeatedloading, and evaluation of the performance of a granular
pavement constructed using RCA and subjected to real trafficloading. As the first step of such a detailed investigation of
RCA, this paper presents the results of classification tests ofsix different RCA samples to analyze the feasibility of using
them in unbound granular pavement constructions comparing
their classification properties with those of the conventional
unbound granular materials used in Queensland.
2. MATERIALS USED
For this investigation, two main commercially available RCA
products, namely RM001 and RM003, obtained from aleading concrete recycling plant in Queensland were utilized.
Material sources are demolished building (slabs, floors,
columns and foundations), bridge supports, airport runways
and concrete road beds. The collected materials areundergone specified crushing process to produce RM 001 and
RM 003. Table 1 shows the maximum percentages of theconstituents that can be consisted in RM001 and RM003 at
the plant output. These two materials were blended indifferent percentages by weight to form another four samples
to represent various combinations of constituents. New
sample types with their blending percentages are showing in
Table 2.

575

RM 001

100--

RM 003

100

20

15

approximately half of sand and half of gravel with slightly
susceptible to gravel.
100

Percent Passing %Table 1: Percentage limits of constituents of two main
RCAmaterials.
Recycled
Maximum Limit of each Constituent
Material(Percentage by mass)Type
Reclaimed
*RAP
Brick
Concrete*RAP – Reclaimed Asphalt Pavement

RM1-100/RM3-0

90

RM1-80/RM3-20

80

RM1-40/RM3-60

RM1-60/RM3-40

70

RM1-20/RM3-80

60

Mat.Type2.1-min

RM1-0/RM3-100
Mat.Type2.1-max

50
40
30
20

Table 2. New RCA samples with blending percentages
Material name
Mixing percentages by mass (%)
RM001
RM003
RM1-100/RM3-0
1000
RM1-80/RM3-20
80
20
RM1-60/RM3-40
60
40
RM1-40/RM3-60
4060
RM1-20/RM3-80
2080
RM1-0/RM3-100
0
100

10
0
0.01

0.1

1

10

100

Grain size-mm

Fig 1. Gradation curves for six samples and maximum &minimum curves of subtype 2.1 materials

3. LABORATORY TESTING, RESULTS ANDDISCUSSION
Usually RCAs are highly heterogeneous and consist indifferent amounts of impurities and their quantities are notsteady. This makes RCA to have inconsistent classification
properties. Therefore it is essential to characterize theproperties of RCAs through classification tests such as sieve
analysis test, atterburg limits test, proctor compaction test andCalifornia bearing ratio (CBR) to investigate the possible
range of variation of classification parameters. The results
were compared with the specifications of base and subbase
materials of Department of Main Roads in Queensland.
3.1 Grain Size Distribution
Particle size distribution of a particular pavement aggregate
type affects its compressibility, permeability, density...etc [9]. Grain size distribution curves of all six testing RCA samples
obtained from sieve analyses are shown in Fig 1. Further the
maximum and minimum grading curves of material subtype
2.1 have been drawn in same figure. Material subtype 2.1 isapplied for base layers by Department of Main Roads,
Queensland and its gradation curves are drawn as their
specifications. These demarcating lines are showing that finecontent of the six samples are within required range but close
to the minimum curve, while the course particles haveexceeded the maximum gradation curve at the top. But all thecurves show that a greater fraction of their curves are laid
within the specified range.
According to the unified soil classification system (USCS),
except ‘RM1-0/RM3-100’, other five samples are categorizedas ‘well graded gravel’. The ‘RM1-0/RM3-100’ present

Table 3. Plasticity Index of two main materials
Fine Sample RM1-100/RM3-0 RM1-0/RM3-100
LL
21
27
PL
15.6
20
PI
5.4
7
3.2 Plasticity
Cone-penetrometer method was followed to deter-mine theliquid limit (LL) as well as plastic limit test was done forplastic limit (PL) for the sample of fines passing 0.475 mm
sieve.
Fines
of
only
‘RM1-100/RM3-0’
and‘RM1-0/RM3-100’ were tested since other four samples are
blended of these two. The LL values and mean values ofplastic limit (PL) test (Atterberg limit test) were summarizedin Table 3. According to the Department of Main Roads
specifications the maximum plasticity index (PI) is 6 formaterial subtype 2.1(See section 3.1). ‘RM1-100/RM3-0’ iswithin that range but ‘RM1-0/RM3-100’having PI of material
subtype 2.3 which is given maximum PI as 8. The subtype 2.3
is applied for upper subbase layers by Main Roads.
3.3 Compaction
The moisture density curve (compaction curve) of a soil is anindicator of the sensitivity of the density with respect to thevariation of moisture content in the materials [10]. Materials
with flat curves can tolerate a greater amount of variations in
the moisture content without compromising much of theachieved density from compaction. In contrast, materials withsharp curves are extremely sensitive to the optimum value
during compaction.

576

Table 5. Variation of CBR values with different moisturehomogenization periods

1.9
1.88

CBR %

1.86
RM1-100/RM3-0

Dry Density g/cm3

1.84

Sample Type

RM1-80/RM3-20

1.82

No
moisturehomogeni
zationperiod3 hrs
moisturehomogeniz
ation period*8 hrsmoisturehomogeni
zationperiod

RM1-100/RM3-0

61

74

55

RM1-80/RM3-20

56

63

-

RM1-60/RM3-40

55

66

-

RM1-40/RM3-60

50

61

-

RM1-20/RM3-80

48

60

-

RM1-0/RM3-100

46

50

53

RM1-60/RM3-40

1.8

RM1-40/RM3-60

1.78

RM1-20/RM3-80

1.76

RM1-0/RM3-100

1.74
1.72
1.7
1.68

5

7

9

11

13

15

17

19

21

Moisture Content %Fig 2. Compaction curves for Six Samples
Table 4. Optimum Moisture Content and Maximum Dry
Density for Six Samples
Sample Type
Optimum
Maximum Dry
Moisture Content density (g/cm3)%
RM1-100/RM3-0

13.2

1.748

RM1-80/RM3-20

13.2

1.768

RM1-60/RM3-40

13.3

1.822

RM1-40/RM3-60

13.5

1.856

RM1-20/RM3-80

14.0

1.836

RM1-0/RM3-100

14.2

1.846

Standard proctor compaction test in accordance with
Australian Standards [11] was performed on each testing
sample and the results are shown in Fig 2. Optimum Moisture
Content (OMC) and Maximum Dry Density (MDD) values ofeach material obtained from Fig. 2 are tabulated in Table 4.
The range of the variation of MDD and OMC are relatively
small as 1.748-1.846 g/cm3 and 13.2-14.2%, respectively. It
can be seen that with the increase in fines contents of amaterial, both OMC and MDD increase as fines can absorb
more water and can reduce the void volume by filling thevoids between larger particles.
Furthermore, with the increase in fine in the samples thecurves are more pointed showing the sensitivity of mixture ofwater. And it is also interesting to note that all the samples are
more sensitive to the moisture variation on the dry side of the
OMC curves than the wet side of the OMC.

*CBR was conducted only for two samples to observe thestrength gaining pattern
3.4 CBR tests
CBR characterization is widely used in pavement industry toprovide a relative measure of strength, elastic modulus andmoisture durability across various road materials forstructural design purposes [9]. CBR tests were performed as
specified in AS 1289.6.1.1 – 1998 [12]on the six samples
compacted at their corresponding OMCs.
In order to find out the optimum moisture homogenization
period (time between material mixing with water andcompaction), a series of CBR tests were performed on RCA
samples
preparedallowing
different
moisturehomogenization periods (e.g. 0, 3, 8 hours) since the main
components of crushed concrete; aggregates, cement motar,
sand need specific time period for uniformly moisture
distribution. The results of these CBR tests are tabulated inTable 5.
The samples under ‘No moisture homogenization period’
have given the lowest CBR values since they did not haveenough time for homogenization of moisture and therefore
lower the compaction and CBR values. 3 hours curing period
showed higher CBR values since the materials have taken
sufficient time for uniformly moisture distribution whichhelps for properly compaction of the samples. But
RM1-100/RM3-0 showed lower CBR values for 8 hours
curing period while RM1-0/RM3-100 had its highest CBRvalue than previous conditions. RM1-0/RM3-100 consists on
more fines, clay bricks, and RAP particles which need moretime and 3 hours insufficient for uniformly moisture
distribution. In the RM1-100/RM3-0 neither clay bricks nor

577

RAP, but when higher the curing time (8 hours) the cement
motar separately harden themselves and the inter particle
connection between those the harden cement motar andrecycled aggregates caused poor compaction as well as poorload transfer through materials under loading. Therefore 3
hours moisture homogenization was followed for the next
CBR test series.
3.4.1CBR test for unsoaked samples
Next CBR test series were conducted to investigate the effects
of unsoaked curing periods of the compacted samples on
CBR values. Each sample was prepared following the 3 hours
curing period and followed standard compaction procedure.
The sample compacted into the mould was cured (unsoaked)
in a sealed container for different periods (e.g. 4 and 8 days)
with 4.5kg surcharged load before testing. The variations of
CBR values with curing periods for each RCA are shown in
Table 6. 4 days curing for compacted samples were followed
since it is followed for standard aggregates by the Departmentof Main Roads. 8 days curing period for compacted samples
were applied to observe the strength gaining with time.
RM1-100/RM3-0 and RM1-80/RM3-20 had slightly lower
values of their CBR after 8 days curing of the compacted
samples. The fine particles of these compacted samples weresettled themselves under 4.5 kg surcharge load for 8 days.
This caused to form little voids between irregular shapes
recycled aggregates since these two sample types have very
low amount of fines comparatively other samples. This makes
weak inter connection in the upper part of the compacted
sample while load transferring and caused for low CBRvalues.

90
80

81

77 76

75

70

Table 6. CBR Values for Different Curing Period ofCompacted Samples
CBR %Sample Type

Mixture-3 hrs
curing &compacted
sample cured
4days

RM1-100/RM3-0

8178

RM1-80/RM3-20

7775

RM1-60/RM3-40

74

75

RM1-40/RM3-60

63

64

RM1-20/RM3-80

60

60

RM1-0/RM3-100

58

41

RM1-60/RM3-40, RM1-40/RM3-60, RM1-20/RM3-80 had
almost same but slightly higher CBR values respectively.
Higher compactability due to the high fine fraction and highwater absorption for cement motar strength gaining affecttheir greater load bearing strength.

74
63

63

60
CBR %

Mixture-3 hrs
curing &compacted
sample cured
8days

60

60 58

58
47

50
40
30
20
10
0

RM1-100/RM3-0

RM1-80/RM3-20

RM1-60/RM3-40

RM1-40/RM3-60

RM1-20/RM3-80

RM1-0/RM3-100

Sample Type
Mixture-3 hrs curing & compacted sample cured 4days
Mixture-3 hrs curing , compacted sample cured 4days, 4days soaked
oaked results
Fig 3. Summary of CBR values for best unsoked and soaked results
578

RM1-0/RM3-100 had its lowest CBR values and this is due tothe lower intrinsic particle strength of crushed bricks andreclaimed asphalts (RAP) which led to lower overall bearing
strength of RCA [13]. Also the poor interlocking system dueto the crushed clay brick and sand decrease the load transfer
capacity of the compacted samples.
3.4.2 CBR test for soaked samples
Testing the CBR values of pavement materials after soaking
is more significant to observe the strength of materials underfully saturated condition. This is highly applicable for floodrisk areas in material selecting and pavement designing. The
Department of Main Roads in Queensland has introduced
minimum required soaked CBR value as 80 for base layers.
The soaked CBR results and the best CBR values which wereshown by the mixture cured 3 hours and compacted samples
cured 4 days are shown in Fig 3. For the soaked test, materials
were cured 3 hours prior to compaction and the compacted
samples also cured for 4 days with 4.5kg surcharged loadprior to testing. It shows that soaked CBR values are slightly
less than the unsoaked values which has the same curing
times but not reached the minimum requirement of the Main
Road’s specifications.
4 CONCLUSION

RCAs have fewer fines but higher percentage of coarse
fraction. The six curves are shown minimum required
particles size distribution of material subtype 2.1.
The proctor compaction test gave relatively higher water
absorption for maximum dry density. Cement, rock dust
and clay bricks cause for higher compaction withincrease the RM 003 portion but lower the density.
However the density varied in a small range giving thelowest value for ‘RM1-100/RM3-0’ which represents
only crushed concrete aggregate.
‘RM1-100/RM3-0’sample
hascertainly
very
appreciable results at CBR test at the both soaked andunsoaked. ‘RM1-80/RM3-20’sample also has comeclose to ‘RM1-100/RM3-0’ with CBR results but lesser
values. However, this improvement is laid in the marginof the minimum requirement of the pavement
aggregates’ CBR values which employ for base layer(CBR 80 for soaked condition, [14]) as introduced byMain Roads in Queensland. But it is possible toconclude that six sample types are showing appropriate
strength for employing as a subbase aggregate as well as
being able to be reusing in base for low or average
traffic pavements.

5 REFERENCES[1] Amnon K. Properties of concrete made with recycled
aggregate from partially hydrated old concrete. Cement
and Concrete Research 2003; 33:703-711.
[2] Kou S-c, Poon C-s, Agrela F. Comparisons of natural andrecycled aggregate concretes prepared with the additionof different mineral admixtures. Cement and Concrete
Composites 2011; 33:788-795.
[3] Poon C S, Shui Z H, Lam L. Effect of microstructure of
ITZ on compressive strength of concrete prepared withrecycled aggregates. Construction and Building
Materials 2004; 18:461-468.
[4] Hansen T C. Recycling and demolished concrete andmasonry. In: Taylor & Francis e-library; 1992.
[5] Chini S A, Sergenian T J, M.armaghani J. Use of recycled
aggregates for pavement. In: Materials for the newmillennium Washington: 2001.
[6] Nataatmadja A, Tan Y L. Resilient Response of recycledconcrete road aggregates. Transportation engineering
2001.
[7] Park T. Application of construction and building debris asbase and subbase materials in rigid pavement. Journal ofTransportation Engineering 2003; 129:558-563.
[8] Poon C S, Chan D. Feasible use of recycled concrete
aggregates and crushed clay brick as unbound roadsub-base. Construction and Building Materials 2006;
20:578-585.
[9] Papagiannakis A T, Masad E A. Pavement design andmaterials. john Wiley & Sons Inc. Hoboken New jersey;
2008.
[10] Poon C-S, Qiao X C, Chan D. The cause and influence ofself-cementing properties of fine recycled concrete
aggregates on the properties of unbound sub-base. Waste
management 2005:1166-1172.
[11] Australia C o S. Methods of testing soils for engineering
purposes. In: Method 5.1.1: Soil compaction and density
tests-determination of the dry density/moisture content
relation of a soil using standard compactive effort. 2003.
[12] Australia C o S. Methods of testing soils for engineering
purposes. In: Method 6.1.1: Soil strength andconsolidation tests-determination of the California
Bearing Ratio of a soil-standard laboratory method for aremoulded specimen. 1998.
[13] Boudlal O, Melbouci B. Study of the behavior of
aggregates demolition by proctor and CBR tests. In:
Material, design, construction, maintenance, and testingof pavement. Changsha, Hunan, China: 2009.
[14] MainRoads Q D o T a. Main roads technical standard:
MRTS35 - Recycled materials for pavements. In: 2010.

ACKNOWLEDGMENT
The authors thank Alex Fraser Group, Queensland forproviding recycled concrete aggregates for this experimental
investigation.
579

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Estimation of consolidation properties of Holocene clays with artificial neuralnetwork
Kazuhiro ODA 1, Minsun Lee 2 and Shotaro Kitamura 3
Osaka University, JapanABSTRACT
It is required to estimate appropriately the spatial distribution of consolidation properties, in order to predict accurately
the consolidation settlement of sea bed ground due to large scale reclamation. It is necessary to estimate the soil
properties at arbitrary position in the ground from the data of soil investigation. In this study, an artificial neuralnetwork was applied to interpolate spatially the consolidation properties, such as natural water content, void ratio,
plastic index, compression index and pre-consolidation pressure. The estimation accuracy of consolidation properties
was judged based on the four indexes, such as R2, G, MARE and SR. The artificial neural network estimates theappropriate consolidation properties with high accuracy, so that the availability of spatial interpolation of consolidation
properties through artificial neural network was confirmed.
Keywords: Neural Network, Geo-informatics, Spatial interpolation, Osaka Bay, Consolidation properties

1. INTRODUCTION
Many man-made islands such as Kansai International Airport,
Port Island, Yumeshima Island et al have been constructed in
the Osaka Bay. In addition, the construction of someman-made islands is planned presently. Man-made islands areusually constructed for a long term, because man-madeislands are very large. On construction of man-made islands,
they are usually divided into some construction of sections.
The construction histories are different in each constructionof section, differential settlement of man-made islands occurs
necessarily. Therefore, the prediction of differential
settlement must be required to control the construction workof man-made islands.
Holocene clays more than 20m, which is under almost
normally consolidated condition, accumulate in Osaka Bay.
The differential settlement of man-made islands in Osaka Bay
is dependent on mainly the consolidation behavior of
Holocene clay layer. The spatial distribution of consolidation
properties, such as compression index, coefficient of
permeability et al, in Holocene clay layer must be elucidated
to predict the differential settlement of man-made islands
accurately. However, it is difficult to estimate the spatial
distribution of consolidation properties from limited numberof soil investigations.
By the way, an artificial neural network is one of theinformation processing systems which have been advanced
recently. This technique has been applied to data mining.
Data mining is the process of discovering new patterns fromlarge number of data sets. That is, locations at which soil
investigations were carried out and geotechnical information
obtained from soil investigations can be chosen as data sets.
The connection between the locations at soil investigations

and geotechnical information could be given throughapplying an artificial neural network. The authors havestudied about the spatial interpolation of soil properties byartificial neural network [1]-[3]. Consequently, the artificial
neural network was applied to estimate the soil properties at
an arbitrary position in the ground. That is, the spatial
interpolation of soil properties was carried out through an
artificial neural network.
In this study, the spatial distribution of soil properties, which
are used to predict the differential settlement of man-madeislands, is carried out through an artificial neural network.
That is, an artificial neural network is applied to estimate thesoil properties at an arbitrary position in Holocene clay layer.
The accuracy of estimation to the soil properties is judged
through various statistical indices. Also, the spatial
distribution of soil properties in Holocene clay layer isvisualized. The availability of artificial neural network to
spatial interpolation of soil properties is discussed based on
the accuracies.
2. ARTIFICAIL NEURAL NETWORKAn artificial neural network is an information processing
systems, in which the nerve cells (neurons) in human brain
are mimicked mathematically. Fig. 1 illustrates thearchitecture of artificial neural network used in this study.
This type of artificial neural network is called backpropagation neural network, which has been most popularly
used. The back propagation neural network is consists of the
input-layer, one or more hidden-layer, and the output-layer.
Each layer is connected by neurons. In this study, the soil
properties at an arbitrary position in the ground should be
estimated. Then, the latitude, longitude, and altitude at target
position at which the soil properties should be estimated werecorresponded to items of the input-layer. Further, theconsolidation pressures in the oedometer tests were added to580

investigations were applied to determine the optimumarchitecture of artificial neural network.
North latitude (°)

3.2 Analysis Procedure
East longitude (°)

Soil property

Altitude (m)Input-layer

Hidden-layer

Output-layer

Input items
1. North latitude (°) 2. East longitude (°) 3. Altitude (m)4. Consolidation pressure* (kPa) (*only use for output-item”3”)
Output items
1. Liquid limit (%) 2. Natural water content (%)
3. Void ratio (at each load on standard consolidation test)

Fig. 1. Architecture of artificial neural network

items of the input layer provided that one dimensional
compression curves were estimated. On the other hand,estimated soil properties were corresponded to items of theoutput-layer.
3. ANALYSIS
3.1 Subject Area for Spatial Interpolation
There are large amounts of geotechnical information obtained
from soil investigations in many construction projects in
Kansai area and they have been accumulated into
geo-informatics database, called GIbase [4]. Figure 2(a)shows locations at soil investigations included in GIbase. The
red dots in Fig. 2 denote the locations at which soil
investigations were carried out. Over 25,000 boring
investigations in Kansai Area is accumulated into GIbase. In
this study, the spatial interpolation of soil properties in
Holocene clay (Ma13) in Kobe Airport in Osaka bay was
carried out. Figure 2(b) shows the locations at soil
investigation in the subject area of this study. 99 boring

The architecture of artificial neural network must be
determined to estimate optimally the soil properties at anarbitrary position in the ground. The determination process of
the architecture is called as “Training”. A large number ofpairs of the input and output items is required in “Training”
process. The method of “Training” to the artificial neuralnetwork is the following. Firstly, the initial network has
randomly assigned weights. When input data are given to
input-layer of this network, they are propagated through thehidden-layer to output-layer, and then output data can beobtained. However, there are differences between theestimated output values and real values. Then, the network
and weights are updated automatically to minimize the errors.
“Training” is repeated until the repetition reaches
preliminarily given number. The optimal network structure isdetermined in this way.
3.3 Indexes for judging Optimum
The following four indexes, R2, G, MARE and SR, were used
to choose the optimum architecture of artificial neuralnetwork.
R2 (coefficient of correlation) is popularly used in statistics
for in a liner regression between estimated value and target
one. It is defined as follows:

R2 

1 N o
  xi  x o  xip  x p 
N i 12
1 N o
  xi  x o 
N i 12
1 N p
  xi  x p 
N i 1

where N number of pair of samples, xi value of each sample,
x mean value of samples. Also, superscripts o and p denoteestimated value and target value, respectively.
G (prediction accuracy) is defined as follows [5].

Kobe Airport Island

Kobe Airport

Kansai International Airport

Fig. 2(a). Locations at soil investigation in GIbase.

Fig. 2(b) Locations of boring investigation for subject area

581

(1)

 RMSE 
G  1   100
RMS 


(2)

4.1 Natural Water Content

where RMSE (root mean squared error) and RMS (root meansquared) are defined respectively, as follows:
RMSE 

2
1 N p
xi  xio 
N i 1(3)

1 N o2
 xi
N i 1RMS (4)The higher the estimation accuracy is, the higher both R2 andG is. MARE (mean absolute relative error) is defined asfollows:
MARE 

1 N p
 xi  xio
N i 1(5)

The higher the estimation accuracy is, the lower MARE is.
Also, both G and MARE are indexes to judge the estimation
accuracy to a set of estimated values. On the other hand, SR(success rates) is used to judge the accuracy of each estimatedvalue. It is based on an absolute error r at each estimated value.
ri is defined as follows:ri 

4. ESTIMATION ACCURACY

xip  xio

(6)

xio

In this study, in the judgment of the estimation accuracy,
three different ranges of SR (%) were used with ri <10%, withri between 10% and 15% and with ri >15% [6].

Table 1 shows the estimation accuracy in natural watercontent. In each case, the value of R2 is over 0.9, the value of
G is almost 80% and the value of MARE is less than 12%, sothat natural water content could be estimated throughartificial neural network. Especially, in WN-1, G is highest
and MARE is lowest. Also, SR with r<10% is highest. It wasconsidered that the estimation accuracy in WN-1 is highest in
the four cases. The artificial neural network could reproduce
the distribution of natural water content.
Fig. 3 shows the 3-dimensional distribution of natural watercontent. The natural water content in Fig. 3 was estimated by
WN-1. Firstly, the subject area was divided into about 67,000
regions. Secondary, the latitude, longitude and altitude ineach region were calculated. Finally, the natural watercontent in each region was estimated by applying the latitude,
longitude and altitude to the artificial neural network in WN-1.
It is very easy to recognize visually the distribution of natural
water contents in Fig. 3. The natural water content decreases
in east to west direction. The natural water content at
southeast corner of sea bed surface in the subject area ishighest. It is over 100%. The natural water content is about
90% at northwest corner of sea bed surface in the subject area.
Also, the natural water content decrease in the direction todepth. The natural water content in bottom layer in Holocene
clay is about 50%. The natural water content hardly varies innorth to south direction.

4.2 Void ratio
Table 2 shows the estimation accuracy of void ratio. In eachcase, the value of R2 is over 0.9, the value of G is more than
93% and the value of MARE is only slightly over 5%, so thatvoid ratio could be estimated through artificial neuralnetwork. The artificial neural network in VR-3 can be

Table 1. Estimation accuracy in natural water content
SR (%)

Case

Repetition
Number

R2

G
(%)

MARE
(%)

r<10

10<r<15

r>15

WN-1

5000

0.934

87.10

11.56

80.9

6.6

12.6

WN-2

10000

0.935

87.13

11.63

79.9

6.8

13.3

WN-3

15000

0.936

87.15

11.60

80.7

7.2

12.2

WN-4

20000

0.936

87.16

11.58

79.3

8.9

11.8

Case

Repetition
NumberTable 2. Estimation accuracy in void ratio
R2

G
(%)

MARE
(%)

SR (%)
r<10

10<r<15

r>15

VR-1

5000

0.912

93.46

5.29

89.2

6.7

4.1

VR-2

10000

0.911

93.36

5.21

90.2

5.7

4.1

VR-3

15000

0.908

93.16

5.45

88.1

6.2

5.7

VR-4

20000

0.911

93.49

5.20

88.7

7.2

4.1

582

estimated most accurately in the four cases, judging from SR.
Fig. 4 shows the 3-dimensional distribution of void ratio. The
void ratio in Fig. 4 was estimated by the artificial neuralnetwork in VR-3. The same way for estimating the natural
water contents was applied to make Fig. 4. The void ratio
decreases in east to west direction. The void ratio at southeast
corner of sea bed surface in the subject area is highest. It is
about 3.0. It is about 2.5 at northwest corner of sea bed
surface in the subject area. Also, the void ratio decreases inthe direction to depth. The void ratio in bottom layer in
Holocene clay is about 1.5. The void ratio hardly varies innorth to south direction.
The distribution of void ratio corresponds to that of natural
water content as shown in Fig. 3. The Holocene clay layer ineast side of Kobe airport is thicker than that of the west side.

Also, both natural water content and void ratio in east side of
the airport are higher than that of the west. It was predictedthat the settlement in east side of the Kobe airport is larger
than that in the west side in the construction. The construction
work of the Kobe airport was controlled in consideration of
occurrence of differential settlement.

4.3 Plastic index
Table 3 shows the estimation accuracy of plastic index. In
each case, the value of R2 is almost 0.94, the value of G isalmost 90% and the value of MARE is about 8%, so thatplastic index could be estimated through artificial neuralnetwork.
Fig. 5 shows the 3-dimensional distribution of plastic index.
The plastic index in Fig. 5 was estimated by the artificial
neural network in IP-1. The same way for estimating thenatural water contents and void ratio was applied to make Fig.
5. The plastic index decreases in east to west direction in thesame way as the natural water content and void ratio. The
plastic index at southeast corner of sea bed surface in thesubject area is highest. The plastic index is lowest at thenorthwest corner of sea bed surface in the subject area. Also,
the plastic index decrease in the direction to depth. Especially,
the variation of plastic index around both surface and bottom
is significant. The void ratio hardly varies in north to south
direction.

4.4 Compression index
Fig. 3. 3-dimensional distribution of natural water content(WN-1)

Table 4 shows the estimation accuracy of compression index.
In each case, the value of R2 is over 0.8. This value is lower

Fig. 4. 3-dimensional distribution of void ratio (VR-3)

Fig. 5. 3-dimensional distribution of plastic index (IP-1)

Table 3. Estimation accuracy in plastic index

Case

Repetition
Number

R2

G
(%)

MARE
(%)

IP-1
IP-2
IP-3
IP-4

5000
10000
15000
20000

0.934
0.935
0.936
0.936

90.31
89.84
89.61
90.12

7.67
7.78
8.09
7.79

583

SR (%)
r<10
72.3
68.7
71.3
72.8

10<r<15

r>15

12.3
14.4
12.3
11.3

15.4
16.9
16.4
15.9

east to west direction.

than that in natural water content, void ratio and plastic index.
The value of G is almost 87% and the value of MARE is
almost 10%. The estimation accuracy in compression index isslightly lower than that in natural water content, void ratio
and plastic index. The artificial neural network in CC-3 canbe estimated most accurately in the four cases, judging fromall four indices for judging optimum.
Fig. 6 shows the 3-dimensional distribution of compression
index. The compression index in Fig. 6 was estimated by theartificial neural network in CC-2. The same way for
estimating the other soil properties was applied to make Fig. 6.
The compression index in upper and middle part of Holocene
clay layer is almost 0.9. It slightly decreases in east to west
direction. However, its variation is extremely little. The
compression index in lower part increases in the direction todepth. It decreases significantly in bottom part of Holocene
clay layer. The maximum of compression index in lower partin the east side of the subject area is greater. That decrease in

Table 5 shows the estimation accuracy of pre-consolidationpressure. In each case, the value of R2 is over 0.95, the valueof G is more than 87% and the value of MARE is almost 12%.
However, SR with r<10% is about 53% to 61%. Also, SR with
r>15% is about 24% to 27%. It is implied that estimation
value with errors less than 10% is made up about 53% to 61%
of the total and that with errors greater than 15% is made up
about 24% to 27%. The artificial neural network in PC-4 canbe estimated most accurately in the four cases, judging from
SR.
Fig. 7 shows the 3-dimensional distribution ofpre-consolidation pressure. The pre-consolidation pressure in
Fig. 7 was estimated by the artificial neural network in PC-4.
The same way for estimating the other soil properties wasapplied to make Fig. 7. The pre-consolidation pressure

Fig. 6. 3-dimensional distribution of compression index(CC -2)

Fig. 7. 3-dimensional distribution of pre-consolidationpressure (PC-1)

Case
CI-1
CI-2
CI-3
CI-4

4.5 Pre-consolidation pressure

Table 4. Estimation accuracy (compression index)
Repetition
G
MARE
R2
Number
(%)
(%)
r<10

5000
10000
15000
20000

0.832
0.852
0.848
0.813

87.10
87.80
87.55
86.49

10.48
10.23
10.44
11.09

60.3
67.0
62.9
63.9

SR (%)
10<r<15
20.6
16.0
17.5
19.1

r>15

19.1
17.0
19.6
17.0

Table 5. Estimation accuracy (pre-consolidation pressure)

Case

Repetition
Number

R2

G
(%)

MARE
(%)

PC-1
PC-2
PC-3
PC-4

5000
10000
15000
20000

0.963
0.964
0.962
0.964

87.33
87.67
87.38
87.66

12.74
12.35
12.21
12.06

484

SR (%)
r<10
52.6
53.6
56.2
60.8

10<r<15

r>15

20.1
20.1
19.1
14.9

27.3
26.3
24.7
24.2

increases in the direction to depth. There are hardly
distinguishing spatial variation in the subject area.

5 CONCLUSION
In this study, the spatial interpolation of consolidation
properties in Holocene clay in Kobe Airport was carried out
through artificial neural network. Main conclusions are
summarized as follows:
1. An artificial neural network could be applied to estimate
the consolidation properties at an arbitrary position in
Holocene clay layer in Osaka Bay.
2. The estimation accuracy in consolidation properties by an
artificial neural network was high. Especially, theestimation accuracy in natural water content, void ratio
and plastic index was excellent.
3. The spatial interpolation of consolidation properties could
be carried out through an artificial neural network.
4. The consolidation properties interpolated spatially could
be visualized three dimensionally, so that it is very easy
to recognize the distribution of consolidation properties.

6 REFERENCES[1] Oda, K, Ikeda, S and Tokida, K, “Spatial interpolation ofsoil properties in Osaka Bay area through artificial neuralnetwork,” Proc Symposium on Geoinformatics andZoning for Hazard Mapping, 2009, pp. 178-181.
[2] Oda, K, “Spatial Interpolation of soil properties in northarea of Osaka bay through artificial Neural Network,”
Advances in Ground Technology & Geo-Information,
2011, pp. 143-150.
[3] Oda, K, Kitamura, S, Lee, MS, “Applicability of
Artificial Neural Network to Spatial Interpolation of Soil
Properties in Kansai International Airport,” Proceedingsof the Twenty-second (2012) International Offshore andPolar Engineering Conference, 2012, pp. 583-586.
[4] Tanaka, Y, and Tsukada, Y, “Geotechnical Database of
Kansai – History and Developments,” Proc Symposium
on Geoinformatics and Zoning for Hazard Mapping,
2009, pp. 85-92.
[5] Juang, CH, Jiang, T and Christopher, RA,“Three-dimensional site characterization: neural network
approach,” Geotechnique, Vol. 51, No. 9, 2001, pp.
799-809.
[6] Erdal, U, Mustafa, L, Ahmet, S and Bekir, KH,“Prediction of lateral effective stresses in sand using
artificial neural network,” Soils & Foundations, Vol 48,
No. 2, 2008, pp. 141-153.

585

Measurement of The Crack Displacement Using Digital Photogrammetry for
Evaluation of The Soundness of Tunnels
A.Kanazawa1, S.Nishiyama2, T.Yano3
Department of Urban Management, Graduate school of Engineering, Kyoto University, Japan
T.Kikuchi4
J-POWER Design Co.,Ltd., Japan
ABSTRACTThe purpose of this study is to develop a measurement technique for crack deformation monitoring by using thedigital image-processing. In the measurement technique proposed here, reflective targets are established at themeasurement points on both sides of the crack as a gauge and the two dimensional displacement of the crack is
calculated based on the coordinates of targets. In this paper, experiments were performed to verify the accuracy
and precision of the measurement and the results showed that the changes of the tensile and shear displacement of
the crack we could measure were found to be about 20μm by using the image taken 1m away from the crack. This
paper also presents the actual tunnel measurement results to demonstrate that the cracks on the tunnel wall can be
identified through simple measurements. After the regular measurements through a year, we found that thetendency of crack deformation and that of slide movement were in close agreement.
Keywords: photgrammetry, image processing, crack displacement, tunnel

1.

INTRODUCTION

In Japan, many infrastructures have been constructedduring the rapid economic growth period in 1960s, andmoreover, Japan often suffers natural disasters such as
earthquakes and typhoons. One of the most challenging
issues in Japan is to know whether these old infrastructuresare safe or not, and the inspection to evaluate theirsoundness will play an important role for countermeasuresuch as reinforcement, repair and reconstruction. In
infrastructures consisted of concrete such as bridges andtunnels, width and patterns of cracks which occur on
surfaces of the infrastructures are one of the mostimportant signs to estimate their soundness [1]. The
prediction of the change of the crack width and length andthe prevention of resultant damages require theestablishment of crack behavior monitoring technology.
Generally, vernier caliper and crack gage are usually used
as measurement tools, but these methods heavily depend
on human efforts, which cause nonobjective evaluation [2].
Considering this situation, this paper presents a study on
the application of a measurement technique using digital
image processing and photogrametry to the developmentof a new measurement system for crack monitoring. In themeasurement technique proposed here, the deformation of
the crack is calculated based on the image taken from anarbitrary camera position. This system has the following
advantages.
・Measurement can be carried out only by taking an image
and does not need human skill, knowledge and experience[3].
・The necessary measurement hardware consists of a digital
camera and a PC, which holds measurement cost in very
low level. [4]・Digital photogrammetric measurement is superior toconventional
measurementinobjectivity
andreproducibility [5].
・It is possible to make a database including also thenumber of cracks and the crack length by using a digital
image [6].
The study aims to develop the measurement method of thecrack displacement on the digital image. In our developed
method, reflective targets are established at themeasurement points around the crack as gauges and thetwo dimensional displacement of the crack, which are
tensile and shear displacement, is calculated based on thecoordinates of the targets on an image taken from arbitrary
position. In this paper, experiments were performed toverify the accuracy and precision of the measurement. In
the experiments, we measure the relation between the
distance and angle of photographic positions from targets
and the accuracy and precision of the measurement results.
This paper also presents the measurement results of thecrack displacements in the actual tunnel damaged bylandslide in order to demonstrate that the soundness of theinfrastructure can be identified through simple
measurements.
2.

PRINCIPLE OF CRACK MEASUREMENT

This section explains techniques and principle formeasuring cracks on digital images taken.
2.1 The process of the measurement
We measure the width and length of the crack as follows.
1) Reflective targets shown in Fig.1 were attached586around the crack as gauges. These targets were
designed with glass beads arranged in circles to
induce a strong diffuse reflection of incident light.
There are four circles on each target. The distance of
one pair of circles on both sides of the crack ismeasured and the change of the distance means thedisplacement. The circles on the target also act asmarks for the perspective projection.
2) A digital image of the targets is taken from anarbitrary camera position and camera angle. The
picture of the condition of measuring is shown in
Fig.2. The digital image taken at an arbitrary camera
angle is converted to the image facing the target by
the perspective projection.
3) The 2-dimensional coordinate values of the centroidsof the circles on the image are measured by using theimage processing.
4) The distances between the circles of the targets on
the both sides of the crack indicate the tensile andshear displacements.
The measurement procedure we developed has thefollowing characteristics [7]:-Measurement involves photographing can eliminate theneed for human skill to improve accuracy as inconventional measurement.

2.2 Basic principles of coordinate measurement oftargetImage processing is used to make it easier to identify
centroids of circles taken on the images. The measurement
precision/accuracy in our measurement system is strongly
dependent on the measurement precision/accuracy of
2-dimensional coordinates of centroids of the circles on thetargets as measurement points. Thus, to improve this2-dimensional measurement precision/accuracy in thepresent study, reflective targets that consist of garranged incircles were established at the measurement points. The
glass beads can induce a strong diffuse reflection of light.
The digital imageries are categorized into gauge imageries
in white and other areas in black through binarization by athreshold of intensity value. The 2-dimensional
coordinates of the centroids are obtained by calculating thecenter of mass of white areas for each circle [8].
Assuming the x- and y-coordinates are x=1~n and y=1~m
in the image coordinate system, the coordinate values (x,y)
of the gauge imageries are calculated by using thefollowing equation. In other words, the coordinates ofcentroids of the circles on the images are obtained by
calculating the center of mass of a particular imagery’s
gauge.
nx = x0 + a xm∑∑ (q(i, j ) × x )ij

i =1 j =1
n

m

(1)

∑∑ q(i, j )50mm

i =1 j =1n

y = y0 + a ym∑∑ (q(i, j ) × y )ij

i =1 j =1nm∑∑ q(i, j )(2)

i =1 j =1

where, x 0 and y 0 are the origin of the image coordinate
system, a x and a y are the pixel size, q(i, j) is the intensity
value of the pixel (i,j).

Fig.1 Reflective targets are established on both sides ofcrack.
-The necessary measurement hardware consists of adigital camera and a PC. That’s why this measurement
system brings down measurement cost.

Fig.3 Measurement point means a circle of the reflective
targets.

587

substituted for (x’,y’) in Eq.(3)&(4) and unknown b i is
calculated. In practice, unknown b i is obtained by theleast-squares method [11].

Fig.4 The image L’ seen from the front view of target is
calculated by this perspective projection.
Next, the image seen from the front view of the targets arecalculated using the the perspective projection [9]. The
perspective projection is based on the collinearity
condition, which means that a measurement point, thecamera and the measurement point’s imagery appearing on
the image taken can be connected by a straight line [10].
The concept of collinearity condition is shown in Fig.3.
The Eq.(3)&(4) show the relation between the coordinates(x, y) in the original image coordinate system and thecoordinates (x’,y’) in the image coordinate system seenfrom the front view of the targets (see Fig.4). There are 8
unknowns in total, namely b i (i = 1~8). Because Eq.(3)&(4) are established for one known point, 4 or more known
points are required to solve the equation. The xy
coordinates of the reflective target is illustrated in Fig.5.

𝑥′ =
𝑦′ =

𝑏1 𝑥 + 𝑏2 𝑦 + 𝑏3
𝑏7 x + 𝑏8 𝑦 + 1

𝑏4 𝑥 + 𝑏5 𝑦 + 𝑏6
𝑏7 𝑥 + 𝑏8 𝑦 + 1

Fig.5 The origin of the coordinate axes is at an arbitrary
point. It is just important to substitute the precise distanceof four centers for (x’,y’) in Eq.(3)&(4).
Finally, the distance between the pair of circles on bothsides of the crack is calculated and the change of the
distance is evaluated as that of crack width and crack
length.
3.

EXPERIMENTS AND DISCUSSION

Experiments were performed to evaluate the accuracy andprecision of the measurement system we developed. The
measurement errors of the coordinates of the centroids
affect the accuracy and precision of the measurement
system. The imagery of target appearing on the image

(3)
(4)

These xy coordinates of centers of four circles are

Fig.6 One target is fixed and another target is displaced
by a micrometer.

Fig.2 A digital image is taken from an arbitrary position.

588

1.20

measured value （mm）

1.00
0.80

0.5m

0.60

1m
1.5m

0.40

one direction movement

0.20
0.00
0.00

0.20

0.40

0.60

0.80

1.00

1.20

displacement of targets （mm）

Fig.7 Points are the relation between the displacement oftargets and the measured value. The values of distance inthe figure mean camera length L.
taken becomes small with an increase of the distance
between photographic position and the target. The
measurement errors of the coordinates of the centroids
depend on the size of the imagery of the target, andincreasing of camera angle also causes an increase of themeasurement errors of the coordinates of the centroids.
That’s why we measured the relation between the accuracy
and precision of the displacement measurement andcamera positions and camera angle. All of images weretaken with the digital camera of 13 million pixels.
To verify the measurement accuracy and precision, areflective target is placed at a fix position and anotherreflective target was attached to a micrometer, and thedigital images were taken at different camera positions andcamera angles. The distance of the micrometer’s artificial
movement D 0 and measured displacement D using theimages taken were obtained to calculate the differences
between the two. Accuracy is defined as the average ofdifference between D 0 and measured displacement D, andprecision is defined as measurement value’s dispersion.
Fig.6 shows the camera positions L and camera angles α inthe experiments.
Fig.7 and Table 1 show the relation between the distanceof the micrometer’s artificial movement D o and measured
displacement D. All images were taken at the camera angle
Table 1 Relation between the accuracy/precision and L(unit: mm) Target size is 50mm by 50mm and focal
length is 50mm.

of 0°. The focal length was 50mm. Increasing L caused thedeterioration of accuracy and precision of the measurement
value, but we can measure the displacement with accuracy/
precision of less than 0.02mm when we taken images up to
1m away from targets. Table 2 shows the relation between
the camera angle α and the accuracy and precision. The
images were taken at camera position of 0.36m. At

two directions movement

Fig.8 Upper figure shows that only tensile displacement
occurs, whereas lower figure shows that shear
displacement occurs simultaneously. The measurementsof displacement need to be recognized whether only
tensile displacement occurs or not.
Table 3 Accuracy/precision of measured displacement
x means tensile displacement and y means shear
displacement. (unit: mm)

different camera angles α, when we take images at lessthan 20°, we can measure the displacements with good
accuracy/precision of 0.01mm. Increasing camera angle αcaused the deterioration of accuracy and precision of themeasurement value. When camera angle α was 60°,
accuracy deteriorated by 0.05mm, and precision
deteriorated by 0.01mm
In general, visual inspection using a venier caliper canmeasure only tensile displacement as the change of crack
width. When displacement of 16mm occurs, we cannot
know whether only tensile displacement or not in case ofusing conventional method. The feature of ourmeasurement system used is to measure tensile and shear
displacement respectively shown in Fig.8. Anotherexperiment were performed to verify the accuracy andprecision of measured tensile and shear displacement.
Table 3 shows the accuracy and precision of measured
shear displacement comparing with those of measured
tensile displacement. All images were taken at the camera
angle α was 0° and at the photogrammetric position L was
0.5m. This figure demonstrated that we can measure theshear displacement and the tensile displacement589

Table 2 Relation between the accuracy/precision and α(unit: mm)

simultaneously with high degree of accuracy andprecision.
In the experiments mentioned above, all images were taken
Table 5 Relation between the accuracy/precision and L(unit: mm). Target size is 400mm by 400mm and focal
length is 300mm.

(5m, 7m, 10m, 15m, 20m). Twice size of targets (100mm
by 100mm) and the lens which focal length was 300mm
were used. Table 5 shows the results when L was morethan 25m and eight times size of targets (400mm by
400mm) were used. From these results, increasing L
caused the deterioration of the accuracy/precision, in thesame way the experiment described above revealed. Table1,
4, and 5 are summarized in Fig.8. In three experiments, notonly camera length L but also the target size and focal
length were varied. This is why the x-axis of Fig.8 is not
camera length L but the diameter of circles in images.
Fig.8 shows that the accuracy/precision of measurement isagree with the diameter of circles in images.
4.

within 1.5m from targets. In actual measurements, it can beconsidered that we have to take images more than 1.5m
away from targets. It can be estimated from Table 1 thatthe accuracy/precision is in proportion to the camera
length L. We are able to assume easily that theaccuracy/precision of measurement become worth morewhen camera length L is more than 1.5m. Therefore, twotypes of the bigger targets were used in the further
experiments. One was the twice size and another was theeight times size of the conventional target. Furthermore,
the lens which focal length was 300mm was used. Table 4shows the relation between the accuracy/precision and L

590

EXA
MPLE
OF
MEAS
URE
MENT
IN AN
ACTU
AL
TUNN
EL

Fig11 The values of the change of cracks’ width areplotted. +5.80mm means that the crack displaced in
5.80mm in the tensile direction. (Tensile is plus.)Table0.64 Relation between the accuracy/precision and L (unit:mm). Target size is 100mm by 100mm and focal length is
300mm.
accuracy/presicion (mm)

0.5

0.4

No.2

0.3

slide Aslide B

50m

0.2

No.3
0.1

ground-water level

0
0

20

40

60

80

100

120

140

tunneldiameter of circles in an image (pixel)

No.1

Fig.8 Points are the relation between the diameter ofcircles in an image and the accuracy/precision. Diamondpoints mean the accuracy and square points mean the
precision.
We applied the proposed method to the crack
measurement in the actual tunnel. This tunnel wasconstructed 20 years ago to prevent the occurrence of thelandslide by draining groundwater in the slope. The
measurements of extensometers indicate the movement of
2-3mm per year in the slope. Fig.9 shows the diagram ofthe slope and the tunnel. The existence of two slide planes
were revealed by the records of boring. The track curves
to the left at a point about 50m from the entrance and goesstraight on parallel to the slide A and B. At a point about
280m from the entrance, two tracks diverge; one track
keeps straight on and another track turns at right angles. In
other words, this track crosses the both slide plane A and
B. The picture of this tunnel is shown in Fig.10. Diameterof the tunnel is 2m, and there are more than 300 cracks
caused by landslide in the tunnel. In order to investigate
the soundness of the tunnel quantitatively, we tried toapply our crack measurement system to this tunnel andinvestigated the relation between the effect of slide andthe displacement crack. Targets were placed on 30
locations, and measurement of crack behaviors were
performed per every two month. In this paper, we report
three examples of the results of the measurements ofcracks. Target No.1 is located on the vertical crack at apoint 15m from the entrance. This area is in the slide plane
A, therefore about 100 vertical cracks occur from a point
0-50m. Target No.2 is also established on the vertical crack
in the branch tunnel at a point 300m. This target is directly
under the slide B, so that there are more than 100 cracks
around target No.2. Last, Target No.3 was placed on thehorizontal crack at a point 100m. The track around target
No.2 is parallel to slide A and B, so this area is difficult tobe affected by the landslide and there are hardly any cracks
around here. Fig.11 shows the three examples of the resultsof the change of crack width measured by our system. It
can be seen from Fig.11 that the targets No.1 and No.2
have displaced at a constant pace for a year. On the other
hand, target No.3 has hardly displaced for a year. From thisresult, it can be concluded that cracks in or under the slide
plane have displaced and cracks which are not under theslide plane have hardly moved. This conclusion suggestthat it is not necessary to monitor all more than 300 cracks

Fig.9 The sectional diagram of the mountain is illustrated
along the branch tunnel which was drilled from thediverging point. Slide A and B are parallel to the tunneland cross the tunnel around the target No.1.

2m

2m

Fig.10 Due to the thin diameter of the tunnel, we wereable to take the digital images of targets less than 0.5m
away from targets.
but to focus on only 10-20 cracks under the slide plane.
With regard to the values of the displacement of targets,the tensile displacement of target No.1 is 5.3mm/year andthat of the target No.2 is 1.8mm/year. These results of thechange of crack width and the result of the measurementsof extensometers described above are in close agreement.
5. CONCLUSIONSIn this study, we developed a measurement technique forcrack deformation monitoring by using a digital camera
image. In this measurement, reflective targets are
established at the measurement points around the crack asgauges and the digital camera image of the targets is
subjected to a process that consists of photogrammetry andimage processing. Experiments were performed to verify
the accuracy and precision of the measurement and theresults showed that the changes of the tensile and shear
displacement of the crack we could measure were found to
be about 20μm by using the image taken 1m away from591

the crack. We also present the actual tunnel measurement
results to demonstrate that the cracks on the tunnel wall
can be identified through simple measurements.
Measurements through a year revealed that the tendency ofthe displacement of cracks is consistent with the results ofthe measurements of extensometers. This conclusionshows that it is enable to measure only cracks in the slide
plane. This may make us decrease much labor and time of
measurements. We plan to do further research response to
the requirement of actual site with the aim to complete ourmeasurement.
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Nakagawa, K., and Nishimura, K.. “A quantitative
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Engineering, Vol.63, No.1, 2007, pp. 86-96.
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“A fundamental study on automated crack
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[3] Fraser, C.S. “High accuracy ground surface
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[6] Nishiyama, S., Ohnishi, Y., Yano, T., Ryu, M. “A
measurement method of tunnel deformation using
digital photogrammetry,” Journal of GeotechnicalEngineering, Vol.64, No.3, 2008, pp. 598-606.
[7] Hattori, S., Akimoto, Y., Ohnishi, Y., Miura, S.
“Semi-automated tunnel measurement by vision
metrology using coded-targets,” Modern Tunneling
Science and Technology, Vol.1, 2001, pp. 285-288.
[8] Trinder, J.C. “Precision of digital target location,”
Photogrammetric Engineering and Remote Sensing,
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[9] Penna, M.A. “Determining camera parameters fromthe perspective projection of a quadrilateral,” Pattern
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[10] Ohnishi, Y., Nishiyama, S., Yano, T., Matsuyama, H.,
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592

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Numerical modeling of a small scale shallow foundation reinforced by
Soil-Mixing
Anna GRZYB, Mahmoud DHAYBI and Frederic PELLETLaboratory of Civil and Environmental Engineering, Institut National des Sciences Appliquees, Lyon, France
ABSTRACT
Soil improvement for the reinforcement of existing foundations has become an important challenge for civil engineers.
One of the major difficulty is to balance efficiency and cost of a method. The study, presented in this paper, is based on
the concept of using the Soil-Mixing (SM) column as a reinforcement for shallow foundations. The numerical finite
element simulation in ABAQUS allows one to identify the influence and the consequences of the SM method on the
behavior of soil and foundation. Parametric study is carried out to investigate the impact of soil constitutive law (theDrucker-Prager criterion and the Modified Drucker-Prager with cap) on the foundation bearing capacity for two cases:a shallow foundation without reinforcement and a shallow foundation reinforced by a central SM column. Calibrationof the numerical models’ parameters was achieved according to the reduced scale laboratory model.
Keywords: Soil-Mixing, ABAQUS, shallow foundation, Drucker-Prager, small scale model
1. INTRODUCTION
The increasing densification of constructions in urban areaand development of transport network creates new challenges
for civil engineers. Soils, which used to be classified as
inappropriate for construction purposes, need to be redefined
because of the economical reasons. Renovations of existingbuildings, which usually lead to an additional loadingtransferred to the foundations, require foundations’ support.
One of the major difficulties is to balance efficiency and costof a method. Up to now many reinforcement methods havebeen developed. Some of the most common techniques are:pre-fabricated vertical drains, preloading, jet-grouting
columns, geosynthetics, light-weight fill, vertical rigid
inclusions (concrete piles, stone columns), Soil-Mixing
method (known also as Deep Mixing), etc.
The objective of this paper is to present results of Finite
Element Method (FEM) numerical simulation, of a small
scale shallow foundation reinforced by the Soil-Mixing (SM)technique. The parametric study leads to investigate theinfluence of soil constitutive law on the foundation bearing
capacity. The axisymmetric analysis of the homogeneous soil
layer is conducted with elastic perfectly plastic constitutive
model with two failure criterions: the Drucker-Prager and theModified Drucker-Prager with cap. The study takes into
account the effect of cap parameters such as eccentricity, its
location and evolution and transition surface radius. Results
are compared with the experimental measurements.
1.1 The Soil-Mixing method
The SM method is frequently applied as a soil improvement,
since 1960s, when it was created in Japan, USA andScandinavian countries [14]. Originally the technique wasdedicated to soils reinforcement but presently it is also used inother geotechnical applications. The major areas ofapplication are: foundation engineering, excavation control,
ground improvement, hydraulic cut-off walls, liquefaction
mitigation, environmental remediation and as a reinforcementof the existing foundations. The technique is able to fulfill
cost-efficiency criteria while being environmental friendly.
The SM method consists in the creation of elements of
mixed-in-place soil with a cementitious material in order tocreate composite stiff elements (Fig. 1).

Fig. 1 The Soil-Mixing column production process [12].
The shape of the SM element and its appliance depend on its
production method. The SM technique classification,
proposed in the Federal Highway Administration Report [4],is presented in Fig. 2. The improvement patterns of the SMcolumns can be classified into four categories: single, wall
type, block type and grid type. All the plane layout patterns ofthese improvement modes are shown in Fig. 3. In this study
behavior of the single SM column has been analyzed. Until
now, no specific algorithm for designers has been created.
Therefore, the investigation by field loading tests isnecessary. However, this type of measurements increases the
project costs. In this context, it seems that numerical approach
can be complementary or even an alternative solution. Manyexperimental ([3], [7], [8]) and numerical ([2], [4], [7], [11])593studies have been performed to investigate the behavior ofsoil reinforced by the SM method.

convergence problems in flow theory, due to six sharp
corners [10].
2.1 The Drucker-Prager criterion
The criterion proposed by Drucker and Prager in 1952 [6],also known as the Extended von Mises criterion, is widelyused in engineering. The yield function of the linear
Drucker-Prager criterion can be expressed by Eq. (1). WhereJ 2 is the second stress invariant, I 1 is the first stress invariant,α and k are material constants.

F = J 2 − αI 1 − k(1)In ABAQUS [1], the linear Drucker-Prager criterion is
written as Eq. (2) and can be illustrated in p-t plane (Fig. 4)
and in deviator plane (Fig. 5). The p parameter is theequivalent pressure stress, β and d are the friction angle andcohesion respectively and t is defined by Eq. (3).
(2)
Fs = t − p tan β − d3q 1 
1  r  
t = 1 + − 1 −   
2  K  K  q  



(3)

Where q is the von Mises equivalent stress, r is the thirdinvariant of the deviator stress and K is a ratio between theyield stress in triaxial compression and tension. The K
parameter controls the dependency of the yield surface on thevalue of the intermediate principal stress. When the triaxial
tension is equal to the triaxial compression K = 1 and t = q(Fig. 5). It is required that 0.778 ≤ K ≤ 1 to ensure that surface
is convex. The Mohr-Coulomb parameters, cohesion c andfriction angle ϕ, can be converted to the Drucker-Prager
parameters as follows:

Fig. 2 The Soil-Mixing technique classification proposed in
the Federal Highway Administration Report [4].

Fig. 3 The improvement patterns of SM method [7].

6 sin φ3 − sin φ
18c cos φd=3 − sin φtan β =

2. CONSTITUTIVE MODELS
Soils are heterogeneous materials, whose behavior is strongly
dependent on factors like: grain size, mineralogy, structure,
initial stress state etc. Various models and criteria have been
proposed [15] (Mohr-Coulomb, Drucker-Prager, Duncan
Chang, Cam Clay) to describe accurately different aspects ofsoil behavior. Some of them have been also apply in the finite
element modeling for geotechnical engineering applications.
Decision, which constitutive law is the most suitable todescribe behavior of analyzed material, has a significant
influence on the final result. The choice of using one or
another constitutive law depends on: soil type, geotechnicalproblem and, most of the time, the possibilities of estimating
constitutive parameters.
The elastic perfectly plastic model with the Mohr-Coulomb
failure criterion is one of the most commonly used strength
theories in geotechnical analysis. Apart of its advantages,
user needs to deal with some limitations. Firstly, theintermediate principal stress σ 2 has no influence on the yield.
Only the major and the minor principal stresses, σ 1 and σ 3
respectively, are taken into consideration. It leads tounderestimation the yield strength of soil and discrepancy
with experimental evidence. Secondly, an irregular hexagonal
section of the yield cone in deviator plane induces the

(4)
(5)

Fig. 4 The Drucker-Prager yield surface in p-t plane [1].
2.2 The Modified Drucker-Prager with cap
The modification added to the model with the Drucker-Prager
criterion allows taking into account effect of the soil
hardening. Moreover, the cap model is able to consider theeffect of stress history, stress path and compaction [1]. The
yield surface of this constitutive law consists of three parts(Fig. 6): a Drucker-Prager shear failure surface, definedpreviously, an elliptical cap, which intersects p axis at a rightangle, Eq. (7), and a smooth transition region between theshear failure surface and the cap, Eq. (8).

594

3 NUMERICAL MODEL
The numerical analysis of the small scale shallow foundation
without and reinforced by a centrally situated SM column
have been carried out to reproduce physical test. The model
parameters are obtained according to the reduced scale (1:10)
laboratory model which consists of a vertically loaded
rectangular foundation, laying centrally on the surface of 1m3
of fine Hostun sand [5]. Two SM columns, the 7 days old andthe 14 days old, have been analyzed. In this research, thesolutions are obtained using the finite element method code
ABAQUS.

Fig. 5 The Drucker-Prager yield surface in deviator plane [1].

3.1 Model geometry and material properties
The shallow foundation bearing capacity has been analyzed
by axisymmetric model. Hence, rectangular rigid foundation
requires to be modeled as a circular one. The equivalentradius, r eq = 130 mm, is calculated according to Eq. (11),where S m is the foundation cross section.

req = 0.5

Fig. 6 The Modified Drucker-Prager with cap model yieldsurfaces [1].
Where R is cap eccentricity (controls the shape of the cap), α
is a parameter use to define a smooth transition region (small
number, typically 0.01-0.05) and p a is an evolution parameterthat represents the volumetric plastic strain driven hardening
or softening, which is defined by Eq. (9). The cap model uses
associated flow (ψ = β) in the cap surface and non-associatedflow (ψ ≠ β) in the shear failure and transition region.
The hardening/softening phenomenon is described as afunction that associates the hydrostatic compression yields
stress p b and the volumetric plastic strain εpl vol , as it isillustrated in Eq. (10). εpl vol | 0 is the initial cap yield surface
position. The hardening curve is obtained from isotropic
consolidation test.
(6)
A = d + p a tan β

(7)

2



α 
 − αAFt = ( p − p a ) + t − A1 − cos β p − Rdpa = b
1 + R tan β
2

(

pl
pl
pb = pb ε vol+ ε vol
0

)

(8)
(9)

(11)

π

The behavior of sand is simulated using the elastic perfectly
plastic model with the Drucker-Prager criterion (DP) and theModified Drucker-Prager with cap (DPC) with parameters
computed from the classical Mohr-Coulomb (MC). The SMcolumn is characterized by model with the Mohr-Coulomb
criterion. Properties of the sand for different constitutive laws
are presented in Table I. The cap parameters used in
calculations are R = 1.2, α = 0.05 and εpl vol | 0 = 0. The
properties of the 7 days old and the 14 days old columns are
given in Table II. Figure 8 shows dimensions and mesh usedin the analysis. The mesh consists of 6-node modified
quadratic axisymmetric triangle elements (CAX6M). In order
to simulate contact between the SM column and soil,interface elements with zero initial thickness with the
Coulomb failure criterion are used. The friction coefficient istaken as tangent of 2/3 of the soil friction angle, μ f = 0.4.

2





Rt
2
 − RA
Fc = ( p − p a ) +α  1 + α − cos β 



4S m

Parameter

Unit

DP / DPCMC

DensityYoung’s modulus
Poisson’s ratio
Friction angle
CohesionYield stress ratioρ [kg/m3]
E [MPa]ν [-]β, ϕ [°]d, c [kPa]
K [-]

1450
50
0.3
52
1.24
1

1450
50
0.3
32
0.2
-

Table I Sand properties.
3.2 Loading and boundary conditions
The numerical model boundary conditions are no horizontal
displacement at the two lateral vertical sides. In the bottom
boundary, displacements are restricted in direction (Fig. 8).
The rigid shallow foundation is modeled by imposed
displacement to the surface, with previously presented
equivalent radius.

(10)
595

4.2 Parametric study of unreinforced foundation
To acquire a better accordance between measurements andnumerical prediction, a more advanced constitutive law has
been used. The Modified Drucker-Prager with cap, presented
above, allows taking into consideration sand’s compaction
hardening. The basic material properties are chosen as for the
preliminary calculation (Table I). The cap parameters
calibration has been achieved according to parametric study
and comparison with measurements. The effects of theparametric study and influence of each parameter on bearing
capacity of unreinforced foundation is presented below.

4 RESULTS
4.1 Preliminary calculation
Preliminary calculation has been carried out for the case of
unreinforced foundation with soil characterized by a modelwith the Drucker-Prager criterion (Table I). In Fig. 9, bothcurves represent the bearing capacity of the vertically loaded
shallow foundation. Curves show good agreement for lower
values of displacement but the final bearing capacity and thefinal displacement are different. As it can be seen, in the firststage of loading, until 6.5 mm, the increase of displacement in
the numerical case is slower than the measured one.
Afterwards, the situation invertes.

Vertical Force [kN]0

Displacement [mm]Fig. 8 Finite element mesh and boundary conditions used innumerical modeling.

4.2.1 Cap eccentricity, R
The influence of different cap eccentricity, R, is presented inFig. 10. Another three cap parameters are assumed as: α =0.05, εpl vol | 0 = 0 and hardening, for analyzed Hostun sand, isobtained from isotropic consolidation test (Fig. 11).
Figure 10 depicts five curves. The best fit curve is obtained
from R = 1.20. As can be seen, in case of R = 0.75 the increaseof displacement is the slowest. Also the lowest value of final
displacement is obtained for this case. The quickest
displacement occurs for R = 2.00. In cases of R = 0.75, R =1.00, R = 1.20, values of borne force are overestimated, in
contrast R = 2.00 case. The more the displacement increases,
the greater the difference between measurements andprediction for R = 2.00 appears.
According to results of the cap eccentricity parametric study,
R = 1.20 has been selected for the following calculations.

Vertical Force [kN]

Displacement [mm]

0

0

2

4

6

8

10

5

2

4

6

8

10

5
10
15
20
25
30
35
40

10

0

Measurements
R = 0.75
R = 1.00
R = 1.20
R = 2.00

Fig. 10 The influence of different cap eccentricity, R, on
bearing capacity of unreinforced foundation.

15
20
25
30

Measurements

35

Drucker-Prager criterion

40

Fig. 9 Comparison between preliminary calculation result andmeasurements.
The final value of the force borne by the foundation is higher
and the displacement domain is larger in the measurements
than in the numerical prediction. It can be explained byinappropriate constitutive model due to the fact that the
failure surface is limited only by shear failure. Since there isno limit imposed on the compaction, numerical prediction isunable to properly reproduce soil behavior.

4.2.2 Transition region parameter, αThe influence of different transition region parameter, α, ispresented in Fig. 12. Cap parameters are assumed aspreviously. In Fig. 12, solid line illustrates the best fit curve
obtained from numerical model, α = 0.050. Other numericalresults correspond to α = 0.000, α = 0.010, α = 0.100,
respectively. The spectrum of numerical results is quite
narrow. The beginnings of all four curves are almost the same
until about 5 mm. Afterwards, the increase of displacement,
for curve corresponds to α = 0.100, starts to be faster than forthe others. The difference between results obtained for α =0.000 and α = 0.010 is negligible. The decrease of α causes anincrease of the final bearing capacity of the foundation.
Based on these results, α = 0.050 has been chosen as the mostappropriate one.

596

described by two hardening functions (Fig. 11). The line withblack markers corresponds to results acquired with Ottawa
sand hardening [9]. Even the first assumed Ottawa sandcharacteristics, reproduces quite well measurements’ curve.
However it is recommended to perform isotropic
consolidation test for better material characterization. The
best fit result has been accomplished by using Hostun sandhardening.

1200

pb [kPa]

1000
800
600
400
Hostun sand

200
0

4.3 Reinforced foundation
Reinforced foundation, has been analyzed with the soil
properties obtained by parametric study. As it has been
mentioned before, the SM column is described by model with
the Mohr-Coulomb criterion. The 7 and the 14 days oldcolumns’ properties are reported in Table II. Figure 14
provides measurements and numerical in terms of verticalforces and vertical displacements.

Ottawa sand

0

0.01

0.02

0.03

Plastic volumetric strain εvolpl[-]Fig. 11 Hardening law of sands used in calculations.
Vertical Force [kN]

Displacement [mm]

0

0

2

4

6

8

10

5
10
15
20

Measurements25

α = 0.00030

α = 0.01035

α = 0.05040

α = 0.100

4

6

8

10

5
10
15
20
25
30
35
40

14 days

2000
1.8
0.2
40
15
700

2000
3.0
0.2
40
15
1100

0

5

Vertical Force [kN]
10

15

20

5

Displacement [mm]Displacement [mm]2

7 daysρ [kg/m3]
E [GPa]
ν [-]
ϕ [°]
ψ [°]
c [kPa]0

Vertical Force [kN]0

Unit

DensityYoung’s modulus
Poisson’s ratio
Friction angle
Dilation angle
Cohesion

Table II Soil-Mixing columns properties.

Fig. 12 The influence of different transition region parameter,α, on bearing capacity of unreinforced foundation.

0

Parameter

Measurements
pb = 25 kPa

10
15
20
25
30
35

pb = 40 kPa

40

Hostun sandWithout reinforcement - measurements
Without reinforcement - numerical prediction
Reinforced 7 days - measurements
Reinforced 7 days - numerical prediction
Reinforced 14 days - measurements
Reinforced 14 days - numerical prediction

Ottawa sandFig. 13 The influence of different hydrostatic compression
yield stress, p b , and hardening on bearing capacity of
unreinforced foundation.
4.2.3 Hydrostatic compression yield stress, pb, andhardening
The last but not least analyzed parameter is hydrostatic
compression yield stress, p b , in relation to material hardening.
As it can be seen in Fig. 13, calculations for four possibilities,
have been carried out. Obtained results can be divided into
two categories: p b is defined as constant or p b is a function
according to Eq. (10). Cases, which belong to firstcategory, p b = 25 kPa and p b = 40kPa, are characterized byearly plastic failure, represented by a sudden drop in bearing
capacity. The higher value of p b , the higher final bearing
capacity of foundation. For the second category, sand isFig. 14 Comparison between numerical predictions andmeasurements SM reinforced foundation.
Numerical prediction of the behavior of the 7 days old SM
column presents a good agreement with measurements.,
despite the fact that predictions overestimate a little bit theforce. Displacement results between 0 and 2 mm have almost
the same values for both columns. Numerical predictions forboth reinforced cases are similar to each other and the final
bearing capacity of the reinforced shallow foundation ishigher. In case of the 14 days old column, predictions
overestimate force for displacements 0 – 3 mm and 12 mm –20 mm, and slightly underestimate between 3 mm and 12.

597

Despite the small differences, obtained results represent well
the behavior of reinforced and unreinforced foundation. The
mentioned differences might be due to the definition of theinterface between SM column and sand or idealized contact
between the foundation and soil layer.
5 CONCLUSIONSThis study was based on the concept of using the SM method
as reinforcement for a shallow foundation. The numericalsimulation was an attempt to identify the influence and theconsequences of the SM method on the behavior of the soil
and the foundation. The finite element axisymmetric analysis
provided good values in terms of the bearing capacity. The
positive effect of the reinforcement on the foundation was
clearly highlighted. The value of the load to be borne by thefoundation is increased significantly, and the displacements
are substantially reduced.
The results lead us to conclude that the more advanced
constitutive law needs to be used. The models we recommend
are models which take into account, not only shear failure of
the soil, but also compaction which is poperly described by acap. Performed parametric study, with the Modified
Drucker-Prager with cap, leads us to the conclusion that themost influential parameter is hydrostatic compression yieldstress, p b which is directly related to hardening phenomena.
The slightly smaller impact on the result is due to cap
eccentricity, R. The transition region parameter, α, seems tobe less significant.
Preliminary prediction evidenced that calculation which only
includes the Drucker-Prager criterion, may be helpful
but insufficient in the design process. Therefore, when it ispossible to analyze soil in laboratory, more advanced
constitutive laws should be used. Taking these paper findings
into account, we can see that results obtained from theadvanced calculations, might be considered on a par with thefield test. However, the necessity of calibration model
parameters according to physical test generates somedifficulties. Therefore, a back-calculation method should be
used to complete this study.

[4] Bruce D., Federal Highway Administration Report, USA
2000.
[5] Dhaybi M., Grzyb A., Trunfio R., Pellet F., “Foundations
reinforced by soil-mixing: Physical and numericalapproach” in Proc. TC 211 IS-GI Brussels, vol. III, 2012,
pp. 137-146.
[6] Drucker D.C., Prager W., “Soil mechanics and plastic
analysis or limit design”, Quarterly of Applied
Mathematics, vol. 10(2), 1952, pp. 157-165.
[7] Fang Z., “Physical and numerical modelling of the soft
soil ground improved by deep cement mixing method”
PhD Thesis, Dept. of Civil & Structural Engineering,
The Hong Kong Polytechnic University, May 2006.
[8] Han, J., Sheth, A.R., Porbaha, A. and Shen, S.L.,“Numerical analysis of embankment stability over deep
mixed foundation” Geotechnical Special Publication
ASCE, vol. 126(2), 2004, pp. 1385-1394.
[9] Helwany S., “Applied Soil Mechanics with ABAQUS
Applications”, J. WILEY & SONS, INC., 2007, pp.
61-67.
[10] Jiang H., Xie J., “A note on the Mohr-Coulomb andDrucker-Prager strength criteria”, Mechanics ResearchCommunications, vol. 38, 2011, pp. 309-314.
[11] Kitazume, M., Okano, K. and Miyajima, S., “Centrifugalmodel tests on failure envelope of column type deep
mixing method improved ground”, Soils andFoundations, vol. 40(4), 2000, pp. 43-55.
[12] LIEBHERR, LB series – Soil mixing. Technical report,
Austria, 2012.
[13] Navin M.P., “Stability of Embankments Founded on Soft
Soil Improved with Deep-Mixing-Method Columns”
PhD Thesis, Faculty of the Virginia Polytechnic Instituteand State University, Aug. 2005.
[14] Porbaha A., “State of the art in deep mixing technology:
part I. Basic concepts and overview” in Proc. ICE Ground Improvement, 1998 , pp. 81 –92.
[15] Ti K.S., Huat B.B.K., Noorzaei J., Jaafar M.S., Sew G.S.,“A Review of Basic Soil Constitutive Models for
Geotechnical Application”, Electronic Journal of
Geotechnical Engineering, vol. 14, 2009, pp. 1-18.

6 ACKNOWLEDGMENT
The authors would like to thank the French Ministry of
Economy, Finances and Industry, the local authorities(Départements 77 and 93) which supported this research
program named RUFEX (Renforcement et Reutilisation desplateformes Ferroviaires et des Fondations Existantes)
7 REFERENCES[1] ABAQUS Theory Manual. Version 6.10, 2010
[2] Archeewa E., Puppala A.J., Saride S., Kalla S.,“Numerical Model Studies of Deep Soil Mixing (DSM)
Column to Mitigate Bridge Approach Settlements” inProc. Geo-Frontiers: Advances in GeotechnicalEngineering, 2011, pp. 3286 – 3295.
[3] Bergado, D.T., Ruenkrairergsa, T., Taesiri, Y. andBalasubramaniam, A.S., “Deep soil mixing used toreduce embankment settlement” in Proc. ICE - Ground
Improvement, 1999, pp. 145-162.
598

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Physical modelling of a small scale shallow foundation reinforced by
Soil-Mixing
Mahmoud DHAYBI and Frederic PELLET
Institut National des Sciences Appliquées, Department of Civil and Environmental Engineering, Lyon, France
ABSTRACT
Soil mixing is a stabilization technique which consists in mixing soil with a stabilizing agent to improve its behavior.
The aim of this research project is to identify the influence of Soil Mixing (SM) and analyze the consequences on thefoundation’s behavior. First, laboratory tests are performed to characterize the SM material. The study of unconfinedcompressive strength and Young modulus of different mixtures shows the effect of curing age and cement content.
Second, a reduced scale model is developed in order to show the influence of SM reinforcement on shallow
foundations. It consists in a rectangular foundation, vertically loaded, and laying on the top of a 1m3 tank filled with
Hostun sand. The effects of age of SM columns and soil density were carefully studied. The difference between the
behavior of unreinforced and reinforced cases is clearly demonstrated by physical modeling.
Keywords: soil mixing, shallow foundation reinforcement, Hostun sand, small scale model, bearing capacity
its properties which have been investigated in different
studies. The addition of few cement percentages was able to

1 INTRODUCTION
The growth of infrastructure in urban areas inmost countries of the world has led to a dramatic rise in land
prices.
Consequently,
areas
with
poor
soil
conditions become more and more attractive for development
and construction. The solution to overcome the difficulties
of construction on poor quality ground is either improving
thecompressible
soilorinstallingdeepfoundations. The foundation techniques such as various
methods of soil improvement are usually more costeffective compared to deep foundations such as driven
piles, bored piles or drilled piers [1]. The basic
concepts of soil improvement, which are, densification,solidification,
dehydration,
strengthening, were developed. However, over the last
three decades, many techniques have been developed andapplied to complicated soil, and it is expected that newtechniques for soil improvement continue to be developed.
While conventional methods of soil strengthening passthrough the construction of rigid foundation elements fortransferring the entire loading to a supporting layer, the SM
method opens the door to a new conception of reinforcing
foundations. It is based on the effort distribution between thecolumn of improved soil and the surrounding soil. The
process of SM sets up a new material whose characteristics
are intermediate between those of concrete and soil. Thisground modification improves soil bearing capacity,
contributes in the reduction of foundation settlements andprevents shear deformation [2].
Several soil improvements by SM are available. Soil
reinforcement with SM columns is one of the most wellknown shapes used as an alternative solution to conventional
deep foundations for buildings [2].
The addition of binder to soil leads to significantly changes inincrease the drying rate of soil, change optimum watercontent and increase compressive strength [3]The main objective of this research project is to highlight theeffect of SM columns and their contribution in increasing
foundation bearing capacity subjected to vertical load. Age ofcolumns and soil density were the studied parameters toinvestigate the load transfer between column and soil.
2 SOIL MIXING CARACTERIZATIONWith the aim to characterize the behavior of SM material,some laboratory tests are performed to determine thecharacteristics of three different mixtures with different
cement content (Table 1).
Mix

Cement content

Ratio C/S

Mix 1
Mix 2
Mix 3

280 Kg/m3
210 Kg/m3
140Kg/m3

0.2
0.15
0.1

Ratio
C/W
0.7
0.55
0.4

Table 1: Mix composition (S: Sand, C: Cement, W: Water)

The soil under study is Hostun dry sand HN 31[4] (fig 1) andcement CEM III/C 32.5N is used as a binder. Sand, cement
and water are mixed together during 10 minutes under arotation velocity of 62.5 RPM. Some characterization
tests
were
performed
toassessthe mixture consistency. Slump tests (2 to 4 cm slump) andflow test (27 to 32 cm spread) show that our mix is of
considerable consistency [5].
Small cylindrical specimens of 45 mm diameter and 90 mm
height were prepared according to a specific procedure[6]. The preparation of specimens should be
performed immediately after the mixing operation. The

599

mixture is placed in cylindrical molds in three layers and thematerial is vibrated manually to allow the evacuationof trapped air. The blending and the casting must becompleted in less than 40 minutes. Produced specimens weresealed with PVC lids and stored vertically in dry bags withpaper towel soaked in water. The temperature must
be maintained at about 20 ° C.

16 cm (fig 3).

Fig 3: SM prismatic specimen for dynamic Young modulus
measurement

Fig1: Particle size distribution curve for Hostun sandUnconfined compression tests were carried out on these SM
specimens at 7, 14, 21 and 28 days in order to evidence theeffect of curing time on specimen’s behavior. In every test,
three specimens of the same mix are tested at the same day.
Results show that specimens with the highest cement content
have higher UCS (Unconfined Compressive Strength) which
can reach 7 MPa. However, specimens of mix 3 have thelowest UCS. It’s even hard to unmold them at day 7 and they
can barely achieve 1.2 MPa at day 28.

Table 2 below shows the variation of resistance and Youngmodulus with age. Results show a good agreement with otherresearch [7], [8]. The range of the UCS lies between 1.5 MPa
for fine sand and 5 MPa for medium sand with binder factor
between 120 and 300 Kg/m3. Dividing UCS at 28 days by
UCS at 7 days, we get a factor of 2.3 which is in agreementwith other studies carried on Fontainebleau sand [9].
Our results show also that dynamic Young modulus is higher
than secant Young modulus by 25-40%. This results matchwith other studies carried on concrete, where difference wasby 37% [10].
Age(days)

Resistance
[MPa]

E 50
tangent[GPa]

E 50
secant
[GPa]

7
14
21
28

1,6
2,6
3,3
3,8

2,2
3,3
3,8
4,3

2,8
4
4,5
4,9

Dynamic
Young
modulus
[GPa]
3.6
5.7
7.1
8.3

Table 2: Evolution of static and dynamic young modulus with age ofmix 2 specimens

3 EXPERIMENTAL SMALL SCALE MODEL

Fig 2: Evolution of Unconfined Compressive Strength (UCS) of SM
specimens versus curing timeFor 3 mixtures, the slope of the curve of the UCS as afunction of cure time is significant between 7 and 14 days andthen it is reduced. At 14 days the UCS is substantially
increased. Our study was carried on with mix 2 because it
proves a good compromise between resistance andeconomy. On each specimen, three strain gages are
installed to measure axial strain and determine Young's
modulus. Additionally, dynamic elastic modulus of thespecimens is determined by measuring the resonancefrequency on specimens of prismatic shape of 4 cm x 4 cm x

In order to analyze the behavior of shallow foundations
reinforced by SM columns and composite interface
mechanisms, an experimental setup inspired from WANG
[11] has been built. It consists of a tank 2m long, 1m wide and1m high divided into two parts, 1 m3 each, in which a 1/10
scaled model is placed. The tank is filled by layers of Hostun
dry sand 10 cm thick to ensure appropriate homogeneity andgood compaction of the soil mass.
Since the aim of the experiment is to study the behavior ofsoil and not to dimension the structure, a rectangular steel
plate 25 cm x 20 cm is installed on the top surface of the fill
and models the shallow foundation 2.5 m x 2 m. The plate isvertically loaded using a servo- hydraulic jack. A special
metal frame based on a guidance system on rails allows thejack to be moved easily in the horizontal plane, covering theentire upper surface of the tank.

600

Test

Type
of soilAge of column(days)

Numberof tests

Test 1
Test 2
Test 3
Test 4
Test 5

Dense
Dense
Loose
Loose
Dense

7 days14 days
7 days14 days
-

3
3
2
2
2

Soil
density(KN/m3)
15
15
13.6
13.6
15

Table 3: Performed loading tests

Fig 4: Photo of the constructed tank

Setting up SM column is done by sinking a steel tube in soil
(fig 5). Sand in the tube is extracted using vacuum. Then the
mixture prepared in the laboratory is poured into the tube. The
SM column, which is 450 mm in length and 70 mm indiameter, remains in sand for the curing period (7 days, 14days).

3.1 Loading tests on dense soil
Figure 6 shows the evolution of the bearing capacity of thefoundation subjected to increasing vertical displacement intest 1. After 7 days, foundation and columns exhibit almost
the same behavior. For reinforced cases, vertical stressincreases very sharply, causing the pressure to rise
considerably after 4-5 mm vertical displacement.
Curve for this reinforced case can be divided into three parts.
They start with a sharp increase of axial force (0-5 mm), theslope decreases later on (5-12.5 mm). The third part of thecurves corresponds to the lower slope.
To verify the total force measured by the load cell, force
resulting from the pressure on the soil and force measured on
the column head were added. A little difference exists
possibly due to presence of pressure cells at the bottom of thesteel plate which could lead to the creation of rigid corners.
Thus, pressure measured by soil pressure cells is a little bit
overestimating the real pressure under the steel plate.

Fig 5: Procedure of setting up-mixing of soil columns

The instrumentation of the tank consists in different sensors.
A 100 kN load cell is placed between the plate and the jack tomeasure the vertical force. Two stress sensors are installed
between the plate and the soil to measure the pressure borneby the soil. A small force sensor is placed between the head of
the SM column and the steel plate to measure the pressure
applied on the reinforcement column. Three displacement
sensors are installed on the plate to check that it remains
horizontal.
In order to determine the effect of column age and soil density
12 loading tests were performed (table 3). In these tests, soil
density varies from 15 kN/m3 to 13.6 kN/m3. In the first case,each layer is compacted using a rectangular heavy steel plate
to make the soil denser and more homogeneous. In the second
case, the filling process is carried without any compaction. In
both cases, loading tests were performed for column age 7
and 14 days.

Fig 6: Loading test 1 – dense soil mass

Figure 7 shows the evolution of pressure under the foundation
with vertical displacement.
Stress supported by soil
corresponds to almost 60% of the total stress. For a 25 mm
displacement, stress applied on soil under the foundation isaround 0.2 MPa while 0.32 MPa is the original stress appliedon the steel plate.

601

Fig 7: Total stress applied on the foundation- stress registered by soilpressure cells.

less than 3 mm even after 7 days. In terms of higher loading,the difference increases more significantly and thecontribution of reinforcement also increases.
For 7 kN, the unreinforced foundation resulted in adisplacement of 22 mm, whereas reinforced ones do not
exceed 4 mm.
Taking recent findings into account, we can see thatreinforcing foundations with soil mixing columns leads toincreasing the bearing capacity of foundations and greatly
reduces the extent of settlements.
Contribution of SM columns to support the applied loading
was also investigated (Fig 9). Contribution of SM column andsoil are the ratio between the vertical loading supported by theelement (soil or SM columns) and the total vertical applied
load. Figure 9 shows that SM column takes around 45% of
total charge applied on the foundation. The presence of a stiff
element in contact with the foundation gives to the foundationa higher bearing capacity because it takes an important part ofthe charge.
On the other hand, the same analysis was carried on
reinforced foundation at day 14. It shows that SM columns
take 55% of total applied charge. With aging, SM columns
become stiffer due to cement curing, which makes their
Young modulus higher and makes the stiffness difference
between soil and column larger.

Fig 8: Loading test results carried on dense sand- reinforced andunreinforced cases

Figure 8 shows results of loading tests carried on dense sand.
Test 1, 2 and 5 were repeated to verify the good
reproducibility of the tests. Regarding the results shown
above where soil mass density is 15kN/m3; there is strong
evidence that SM columns have a positive effect on
foundation behavior. After reinforcement, the bearing
capacity of the foundation increased more than twofold after
7 days and almost threefold after 14 days. Reinforced
foundations with different column age have a similar
behavior, but with age, foundation becomes stiffer.
Settlement of the foundations is another aspect which has tobe taken into account. It was reduced after reinforcement
even for low loading values. As shown for a loading of 4 kN,
the unreinforced foundation resulted in a settlement of 8 mm.
On the contrary, reinforced foundations undergo settlements

Fig 9: Contribution of soil and SM column to take charge of theapplied load

3.2 Loading tests on loose soil
In this part, loading tests are performed on loose soil mass
reinforced by SM columns. Composition of columns and theirinstallation process are the same as before. Filling the tank isachieved by layers of 10 cm thick of Hostun dry sand without
any compaction. Sand is properly spread to ensure a good
homogeneity.
In terms of vertical applied load (fig 10), a sharp increase invertical stress takes place at the beginning for a small amountof vertical displacement (0-2mm). This is due to the presenceof a stiff element under the foundation which takes the mostpart of the applied load.

602

After this phase, curve slopes become lower and theverification force is calculated, as explained before. In termsof vertical displacement, reinforced foundations in loose sandpresent much higher vertical displacements in comparison
with dense sand presented before. At the beginning thepresence of SM column gives a sharp increase in bearing
capacity. On the other hand, the tip of SM column is also inloose sand, which makes tip resistance lower.

At the beginning of the loading test, soil is loose and SM
column is the stiff element underneath the foundation which
explains his high contribution. Increasing displacements
leads soil compaction . As a result, soil under the foundation
becomes denser which explains the increase in soil
contribution. The system soil-column is plastified whendisplacements are higher than 4 mm.

Fig 12: Contribution of soil and SM column to take charge of theapplied load

Fig 10: Loading test 3 –loose soil mass

Fig 11: Total stress applied on the foundation- stress registered bysoil pressure cells.

In terms of stresses, pressure cells were also installed between
foundation and soil (Fig 11). It can be seen, these cells record
almost the same pressure. At the beginning and for small
vertical displacements, stress on soil is between 30 and 45%of the total vertical stress. Higher displacements results
higher stress on soil that becomes between 55 and 60% of thetotal applied stress.
Figure 12 shows the contribution of soil and SM columns. For
small displacements (0-6mm), column takes the higher part ofthe vertical applied load. Later on, soil and SM column havethe same contribution, and total load is shared between these
two elements.

Fig 13: Loading test results carried on dense sand, loose sand andunreinforced dense sandFigure 13 shows average results of performed tests (table 3).
Curve from unreinforced foundation in dense sand is similar
to the reinforced foundation in loose sand loaded at day 7.
This similarity shows the importance of the reinforcement
since we could have the same behavior of a dense soil with a
stiffer behavior at the beginning of vertical displacement.
Difference between bearing capacity at 7 and 14 days in bothcases (loose and dense sand) is around 23 %. This differenceis due to cement hydration which makes column stiffer andable to take higher load.

603

4 CONCLUSION
The laboratory investigation on the influence of age andcement content of soil mixing specimens provided valuable
results in terms of resistance, Young modulus andconsistency.
The physical model tests highlight some major differences inthe behavior of reinforced and unreinforced foundations.
With reinforcement, bearing capacity becomes higher andfoundations undergo lower settlements.
Age of SM columns has also an important effect on
foundation’s bearing capacity which is 23% higher at day 14
than at day 7.
In dense soil mass, SM columns contribute to take 45% of thetotal vertical loading at day 7 and 55% at day 14. In loose soil
mass, SM columns contribute to take almost half of the total
vertical loading at day 7 and 55% at day 14.
As a result, SM columns take an important part of the loading
applied on the foundation, which increase its bearing capacity
and reduce stress applied on soil.
Reinforcing loose soil mass with SM columns, makes thefoundation behaves as if it lays on a dense soil mass and evenbetter at small vertical displacements.
This study will be extended by additional tests. Other
reinforcement models have to be tested, especially toemphasise the role of an intermediate granular layer between
the footing and the column. Heterogeneous soil, where deeper
soil layers are denser than the upper ones will also be studied.

[7] Porbaha A., “State of the art in deep mixing technology:
part III. Geometrial characterization” in Proc. ICE Ground Improvement, 2000, pp. 91 –110.
[8] Topolnicki M., “General overview-Deep mixing shortcourse”,
International symposium of ground
improvement , IS-GI Brussels 2012[9] Guimond-Barrett A., “On the strength and durability of
cement-stabilised sands”, in Proc. International
symposium of ground improvement , IS-GI Brussels
2012[10] Zheng L., Sharon Huo X., Yuan Y., “Experimental
investigation on dynamic properties of rubberized
concrete”, J. Construction and building materials, 2008,
pp.937-947.
[11] Wang X., Zheng J., YIN J., “On composite foundation
with different vertical reinforcing elements undervertical loading: a physical model testing study” J. of
Zhejiang university, November 2010, pp. 80-87.
[12] Fioravante V., “Load transfer from a raft to a pile with aninterposed layer”, Géotechnique, pp.121-132.

5 ACKNOWLEDGMENT
The authors would like to thank the French Ministry of
Economy, Finance and Industry, the local authorities(Departements 77 and 93) which supported this research
programe named RUFEX (Renforcement et reUtilisation desplateformes Ferroviaires et des foundations EXistantes).
6 REFERENCES[1] Porbaha A., “State of the art in deep mixing technology:
part I. Basic concepts and overview” in Proc. ICE Ground Improvement, 1998, pp. 81 –92.
[2] Porbaha A., “State of the art in deep mixing technology:
part I. Basic concepts and overview” in Proc. ICE Ground Improvement, 1998, pp. 81 –92.
[3] Sarisseri F., Muhunthan B., “Effect of cement treatment
on geotechnical properties of some Washington statesoils”, J. Engineering geology, 2009, pp.119-125.
[4] Technical
data
sheet
of
Hostunsand http://www.sibelco.fr/item_img/medias/images/ft1
2_hn31.pdf , Last visit June 19th 2012.
[5] AFNOR, “NF EN 206-1: concrete - Part1: specification,
performance, production and conformity”, Freshconcrete, consistence classes.
[6] Guimond-Barrett A., “ Protocole de référence pour laréalisation des mélanges sol-ciment en laboratoire”,
IFSTTAR laboratory, 2011.

604

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Prediction compressive strength of concretes containing silica fume andstyrene-butadiene rubber (SBR) with a mathematical model
A.Haghollahi, M.Shafieyzadeh
Shaid Rajaee Teacher Training University, Iran

ABSTRACTThis paper deals with the interfacial effects of silica fume (sf) and styrene-butadiene rubber (SBR) on workability
and compressive strength of concrete. Analyzing the compressive strength results of 32 concrete mixes performed over
two water-binder ratios (0.35, 0.45), four percentages replacement of silica fume (0%, 5%, 7.5%, 10%) and four
percentages of SBR (0%, 5%, 10%, 15%) were investigated. The results of experiments were shown that in 5% SBR,
compressive strength rises slightly, but when the polymer/binder materials ratio increases, compressive strength of
concrete decreases. A mathematical model base on Abrams’ law has been proposed for evaluation strength ofsilica fume-SBR concretes. This simplified model might serve as useful guides for commixture concrete admixturescontaining of silica fume and SBR. The accuracy of the proposed model is investigated.
Keywords: Concrete, Silica fume, SBR, Compressive strength, Mathematical modeling

1. Introduction

2. Test program and procedure

The strength of concrete is affected by many factors, such as
cement composition and fineness, water-to-cement ratio,
aggregate, age and temperature of curing. There is as yet nosuch formula (mathematical model) that could reproduce theeffects of all these factors adequately in a quantitative
manner, primarily due to a high number of variables [1].
Abrams’ water–cement ratio law in 1918 is still
considered as a milestone in the history of concrete
technology, it is accepted that the largest single factor thatgoverns the strength of concrete is the water to cement ratio.
Originally, concrete was made by mixing cement, aggregates
and water, and uses of admixtures were unknown. The only
cementations material was cement. The present-day,new-generation concretes contain mineral admixtures andlatexes for a variety of reasons. These materials increaseabrasion strength or durability and decrease permeability[2-5], and Abrams’ formulation needs to be modified.
Two different ways of adding polymers to cement
composites have been described [6]:1. Keeping constant the water-to-cement ratio (W/C) toobtain a similar hydration of the cement paste.
2. Fitting the consistency of the composite, by adjusting the
W/C.
In this research, the water to binder ratio is constant (0.35,0.45) considered and the effects of SBR emulsion and silica
fume on fluidity and compressive strength of concrete areinvestigated and a relationship between compressive strengthof concrete with the ratio of polymer, silica fume, water tobinder materials and time of curing in water is proposed.

In this research, two series of samples with water-bindermaterials ratios of 0.45 and 0.35 were prepared. Intend toinvestigation interfacial effects of silica fume and SBR on
compressive strength, experiments base on four levels of
replacement silica fume (0%, 5% , 7.5% , 10% ) and fourlevels of SBR (0%, 5% ,10% , 15%) were performed.
Cubes 150 150 150 mm were cast for compressive
strength test. Before casting, coarse aggregate, sand andmixture of water and silica fume were mixed first. Then,
cement, SBR latex and rest water together with supper
plasticizer were put in the mixer and completely mixed. The
mixed concrete were cast in molds to make specimens, andcompacted by mechanical vibration. The specimens were
subjected to the following three different curing systems:
1) 7-day water curing at 200 c and 53-day dry curing at 200 c(7W53D).
2) 14-day water curing at 200 c and 46-day dry curing at 200 c(14W46D).
3) 28-day water curing at 200 c and 32-day dry curing at 200 c(28W32D).
SBR latex as an aqueous polymer dispersion prefers
dehydration and solidification under the dry conditions [7].
The cement hydration and polymer film in the modified
concretes develop with prolongation cured age, which resultsin enhanced strength. In other words, the combination of the
wet and dry curing is effective for the strength developmentof the polymer-modified concretes. A co-matrix is formed byboth processes [4,8-10]. The specimens were tested forcompressive strength according to BS standard.

605

3. Materials
The materials used in this research, were:
Ordinary Portland cement (Type 1) produced by Tehran
factory, and silica fume, a by-product of the ferrosilicon
Deligan factory. Chemical Properties of cement and silica
fume are given in table 1.
Coarse aggregate with a maximum particle size of 17 mm andfine aggregate with a 3.01 fines modulus were used in theexperiment. The specific gravity and water absorption ofcoarse and fine aggregates were 2.55 and 1.6%, 2.25 and
2.4%, respectively.
A water reducer agent with thecommercial name of Gelenium 110p from BASF was used toadjust the workability of the concrete mixtures. The base ofthis supper- plasticizer is a unique carboxylic ether polymer
with long lateral chains.
The polymer latex used was styrene-butadiene rubber (SBR)
latex with commercial name of Rheomix 141p from BASF.
The properties of SBR latex is presented in table 2.
Table 1- chemical Properties of cement and silica fumeComposition (%)
CementSilica fume

4.

Sio2

21.46

91.7

Al2o3

5.55

1.0

Fe2o3

3.46

1.0

Mgo

1.86

1.8

Cao

63.95

0.9

So3

1.42

0.87

K2o3

0.54

Na2o

0.26

C3S

50.96

C2S

23.1

C3A

8.85

C4AF

10.53

Mean particle
Size (micron)

Butadiene
content

PH

1.01

0.17

40%

10.5

4.3. Effects of silica fume on compressive strength
It is observed a significant improvement in compressive
strength of concrete because of the high pozzolanic natureand void filling ability of silica-fume [14-16]. The
compressive strength of silica-fume-concrete continuously is
increased with respect to reference concrete and reached amaximum value of 7.5% replacement level (Fig. 2). The
effect of silica fume in water –binder 0.35 is less than water–binder 0.45.
4.4. Interfacial effects of silica fume and SBR on compressive
strength of concrete
The compressive strength of concrete is affected by theratio of SBR and silica fume to binder materials. When theratio of polymer/binder is certain, the amount of silica fume
affects strength of concrete. The percentage of silica fumethat optimizes compressive strength remains 7.5% (Fig. 3).
In samples with 5% polymer and 7.5% silica- fume(water to binder ratio 0.45) a decrease in compressive
strength is observed (Fig. 4). The cavitation of superplasticizer induces a decrease in compressive strength ofthese samples. (In making of these samples, the ratio of superplasticizer was kept constant.)
In 15% polymer, with decreasing water to bindermaterials ratio from 0.45 to 0.35, decreasing of compressive
strength is compensated by the reduction of w/b due to theplasticizer effect of SBR. Both phenomena together remaincompressive strength constant.

Table 2 - Properties of SBR

Density(gr/cm3)

In polymer-cement ratio (5%), the compressive strength ofmodified samples was equal and even slightly higher than thatof no polymer samples (Fig. 1). The filling effect of polymer
cause increasing a little of Compressive strength. The
compressive strength is mostly influenced by the bonding
forces generated by hydration reaction of cement [11-13], anda little of increasing is observed. It seems at a p/c ratio 5%, thecontinuity of polymer film is only present through small tiny
bridges on a limited number of spots. Although, the structure
between the polymer film and cement hydrates develops at ap/c of 10%, the compressive strength of concrete decreases incomparison of no polymer concrete. In other words, theinclusion of SBR latex in concrete produces a decrease ofcompressive strength, due to a lower mechanical capacity oflatex with regard to cement paste.

Test results and discussion

4.1. Effects of silica fume and SBR on fluidity of concrete
Silica fume decreases fluidity of concrete but SBR increases
fluidity of concrete. Adding of 15 percent of SBR to samples
with water to binder materials ratio 0.45 induces
self-compacting concrete. The effect of SBR in increasing
fluidity of concrete is more than the decreasing effect of silica
fume in fluidity.
4.2. Effects of SBR on compressive strength606

Fig 4. Effect of polymer-binder ratio on the compressive
strength of polymer modified concrete.

Fig. 1. Effect of polymer-cement ratio on the compressive
strength of concrete(w/c=0.35 and 0% silica fume)(W/b=0.45 and 7.5% silica fume)
4.5. Investigation of main effects and interaction effects offactors in compressive strength

Fig 2. Effect of silica fume- binder ratio on the compressive
strength of concrete (w/b=0.35 and 0% SBR)

The main effect of each factor is shown in Fig 5. This diagramshows means of the compressive strength at various levels ofeach factor. It is observed that decreasing of water to bindermaterial from 0.45 to 0.35 increases compressive strength36.7−29.06
about 26% (× 100) .
Averagely, 7.5% of
29.06
replacement silica fume increseases compressive strength
about 13%. Also adding 15% of SBR decreases compressive
strength about 14%.
A single diagram of interaction effects of factors incompressive strength is shown in Fig 6. Parallel lines in aninteraction diagram indicate there is no interaction effectbetween variables. The greater the departure of lines from theparallel state indicates the higher the degree of interaction, in
this situation, the response at a factor level depends upon thelevels of other factors. For example, between SBR and silica
fume in level of 5% SBR with other levels is observed aninteraction effect. (A decrease in compressive strength ofsamples with 5% SBR and 7.5% silica fume in water to binder
ratio 0.45 is happen.)
Main Effects Plot for compressive
Data Meansw/(c+sf)

p

36.0
34.5
33.0
31.5

Mean

30.0
0.35

0.45

0.00

sf

0.05

0.10

0.15

t

36.0
34.5
33.0
31.5
30.0

Fig 3. Effect of polymer-binder materials ratio on thecompressive strength of polymer modified concrete(w/b=0.35 and 7.5% silica fume)

0.000

0.050

0.075

0.100

7

14

28

Fig 5. Main effects of diagram for compressive strength607

Interaction Plot for compressive
Data Means

0.00

0.05

0.10

0.15

0.000

0.050

0.075

0.100

7

14

28
40
35

w/(c+sf)w/(c+sf)
0.35
0.45

30

35

p
0.00
0.05

30

0.10
0.15

40

p

40
35

sf

30

sf
0.000
0.050
0.075
0.100

t

Fig 6. Interaction effects of factors in compressive strength
5. Mathematical model
The primary factors that affected the compressive strength of
concrete are the ratios of water, SBR, silica fume to bindermaterials and time of curing in water. As per the classicalformulation of Abrams’ law, there exists an inverse
relationship between the compressive strength and water-to
cement ratio of concrete [17,18]. A lot of researchers
introduce relationship between compressive strength and timeof curing in water with a logarithmic equation. Some ofresearchers propose power equations for the effect of silica
fume on compressive strength of concrete [19,20]. But
relationship between compressive strength with all variables
is yet to be well known. In this situation where there is no
prior known relationship between the compressive strength of
concrete with all variables, scatter diagrams are prepared andmeans of values are portrayed in diagrams.
The scatter diagrams for all of the specimens exhibited anonlinear relationship between the compressive strength withthe ratios of silica fume and SBR to binder materials. Fromthe scatter diagrams, it was observed that a two degree
parabolic might be appropriate for showing the effect ofsilica- fume or SBR in compressive strength of concrete. The
relationship of compressive strength with every variable isshown in Figs (7-9). The relationship between compressive
strength with every factor is determined base on regression.

Fig 8. Relationship between compressive strength with Silica
fume to binder materials ratio

Fig 9. Relationship between compressive strength with SBR
to binder materials ratio
The relationship between compressive strength withconsidered variables may be represented by:w

Where FC is the compressive strength (Mpa),
is theb
ratio of water to binder materials, t is time of curing in water(day), s is the ratio of silica fume to binder materials, p is theratio of sbr polymer to binder materials. 𝑎1 , 𝑎2 , 𝑎3 , 𝑎4 , 𝑎5
are constant coefficients. The above equation can be written
the following form:

Fig 7. Relationship between compressive strength and time ofcuring in water

Where a1 , a2 , a3 , a4 , a5 can be determined withmultiple linear regression. The values of these coefficients areshown in below table. The value of the multiple correlation
coefficients (r) has been obtained as 0.95.

608

𝑎1

𝑎2

2.637

𝑎3

0.999

𝑎4

0.98

𝑎5

1.005

0.977

With replace of one to coefficients that are near one, below
equation is obtained.

From above equation can be determined the compressive
strength of concrete containing of silica fume and SBR. To
investigate of the accuracy of above equation, diagrams ofresiduals (the difference of observed and fitted values) for
compressive strength has been drawn (Fig 10). The diagramof residuals for compressive strength shows that the
maximum percent of error for compressive strength is about
10%.
The histogram of residuals has been shown in figure 11.
Although, the distribution residuals doesn’t completely obey
from normal distribution but mean and maximum values ofhistogram of residuals is located near zero.

6. Conclusion
1) The effect of SBR in increasing of fluidity of concrete is
greater than the effect of silica fume in decreasing of fluidityof concrete.
2)In constant of water to binder ratio and combined curingsystem, the compressive strength of concrete in 5% SBR rises
slightly, but when polymer/binder ratio increases, thecompressive strength of concrete decreases.
3) Cement replacement up to 7.5% with silica fume leads toincrease in compressive strength.
4) The decrease of compressive strength is compensated by
the reduction of w/b due to the plasticizer effect of SBR. Bothphenomena together remain compressive strength
approximately constant.
5) The percentage of silica fume that optimize thecompressive strength with adding SBR doesn’t change.
6) The Abrm’s law with some modification is applicable tothe compressive strength of concretes contain of silica fumeand SBR and according to main effects of diagram is
proposed the bellow equation:

Scatterplot of Residual vs compressive
4

The proposed model provides the opportunity to predict thecompressive strength based on the time of curing in water(t),and water, silica fume and SBR to binder materials ratios thatw
we show them with � � , (s) and (p), briefly.

3

Residual

2

b

1
0
-1

7. References

-2
-3
20

25

30

35
compressive

40

45

Fig 10. Diagram of residuals for compressive strength
Histogram of Residual
Normal

20

Mean
StDev

0.0005632
1.170

Frequency

15

10

5

0

-2

-1

0
1
Residual2

3

Fig 11. Histogram of residuals for compressive strength[1] J. Zelic, D. Rusic, R. Krstulovic. A mathematical model for prediction ofcompressive strength in cement–silica fume blends, Cement and Concrete
Research 34 (2004) 2319–2328.
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composites properties, Materials and structures (2005) 467-478.
[4] Ru Wang, Pei-Ming Wang, Xin-Gui Li. physical and mechanicalproperties of styrene-butadiene rubber emulsion modified cement mortars,
cement and concrete Research 35 (2005) 900-906.
[5] S. Bhanjaa, B. Sengupta, Modified water–cement ratio law for silica fumeconcretes, Cement and Concrete Research 33 (2003) 447- 450.
[6] G. Barluenga, F. Herna´ndez-Olivares. SBR latex modified mortar
rheology and mechanical behavior. Cement and Concrete Research 34 (2004)
527–535.
[7] Bing Chen , Juanyu Liu. Mechanical properties of polymer-modified
concretes containing expanded polystyrene beads. Construction and Building
Materials 21 (2007) 7–11.
[8] A. Beeldens, D.Van Gemert, H.Schorn, Y.Ohama and L.Czarnecki. Frommicrostructure to macrostructure : an integrated model of structure formationin polymer-modified concrete. Materials and Structures 38 (2005) 601-607.
[9] LIU jun, XU Chang-wei, ZHU Xiao-yan, WANG Ling-ling,
Modification of high performances of polymer cement concrete. Journal of
wuban university of Technology-Mater vol.18 No.1 (2003).
[10] João Adriano Rossignol, Interfacial interactions in concretes with silica
fume and SBR latex (2009), Construction and Building Materials 23 (2009)817–821.

609

[11] Eui-Hwan Hwang, Young Soo Ko, Comparison of mechanical andphysical properties of SBR-polymer modified mortars using recycled wastematerials.
[12] Ke-Ru Wu, Dong Zhang, Jun-Mei Song, Properties ofpolymer-modified cement mortar using pre-enveloping method, Cement andConcrete Research 32 (2002) 425-429.
[13] Eui-Hwan Hwang, Young Soo Ko, Jong-Ki Jeon, Effect of polymer
cement modifiers on mechanical and physical properties ofpolymer-modified mortar using recycled artifical marble waste fineaggregate, Journal of Industrial and Engineering Chemistry 14 (2008)
265-271.
[14] S. Bhanja, B. Sengupta. (2005). Influence of silica fume on the tensile
strength of concrete , Cement and Concrete Research 35 (2005) 743–747.
[15] H. Katkhuda, B. Hanayneh and N. Shatarat. (2009) . Influence of silica
fume on high strength of light weight concrete, World Academy of Science,
Engineering and Technology 58.
[16] K.C. Biswal and Suresh Chandra Sadangi. ( 2010) . Effect ofsuperplasticizer and Silica Fume on Properties of Concrete. Proc. of Int.
Conf. on Advances in Civil Engineering.
[17] Sandor Popovics , Strength and related properties of concrete aquantitative approach . (1998) John Wiley and sons.
[18] mehta,p.kumar , concrete : Micro structure,properties and materials .
(2006).
[19] S. Bhanjaa, B. Sengupta, Investigations on the compressive strength ofsilica fume concrete using statistical methods, Cement and Concrete
Research 32 (2002) 1391–1394.
[20] Mohammad Iqbal khan, Analytical model for strength prediction of
HPC consisting of cementitious, Architecture civil engineering
environmental (2009).

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Sliding Stability of Dry Masonry Block Retaining Wall Structurewith a Resistance Plate
Akihiro HASHIMOTO1, Yoshio SUEMATSU1, Kazuo FUJITA1, Toshimitsu KOMATSU1 andNoriyuki YASUFUKU1
1)
Department of Civil Engineering, Kyushu University, Fukuoka, Japan
ABSTRACT
There are two types of block masonry in the structural retaining wall, one is "dry masonry" and the other one is "wetmasonry". However, the sliding stability of the dry masonry type retaining wall may be inferior to that in wet masonry
block. Therefore, when focusing on the dry masonry type retaining wall, some arrangements are necessary to increase
the sliding resistance. In this study, a new dry masonry type retaining wall is discussed, which is not only environmental
friendly but also structurally stable. A resistance plate is newly attached to the boundary between upper and lower
blocks to enhance the sliding resistance, which is due to the passive earth pressure mobilized in the plate. In this study,a series of model tests was carried out, paying attention to the position of the resistance plate, mobilized earth pressurein the plate and the horizontal movement of the blocks. It was indicated from the model tests results that the position of
the resistance plate becomes very important to effectively increase the block stability.
Keywords: Sliding resistance, Retaining wall, Dry masonry, Resistance plate, Passive earth pressure
Yo sh io

1. INTRODUCTION
There are two types of retaining block masonries namely "drymasonry" and "wet masonry" to construct the retaining wall.
There is a "dry masonry" and "wet masonry" in theconstruction of retaining wall block masonry. Since, wetmasonry cannot offer certain safety factor against the sliding,a projection at the bottom of basis [1] is given to enhance thesliding resistance on dry masonry retaining wall. In the casethat a certain safety factor against sliding cannot be secured inwet masonry retaining wall, there is a method to enhance thesliding resistance by providing a projection at the bottom ofbasis [1]. However, this method has not been used much inclayey ground because the contact between the projection andthe surrounding ground becomes loose. At the same time theconstruction cost is also relatively higher.
When the effect of projection in the clayey ground is properly
mobilized, the construction of retaining wall can achieve theexcellent safety with minimizing the construction cost. Whenthe effect of projection in the clayey ground can be properly
mobilized, which can be easily constructed at low cost, theconstruction of retaining wall with excellent safety andeconomy becomes possible.
On the other hand, the dry masonry type of retaining wall has
an advantage in the environmental and economic efficiency.
However, it generally resists against the sliding of the block
due to the friction of filling geo-materials, it is considered to
be unstable comparing with the wet masonry retaining wall.
On the other hand, as the retaining wall in dry masonry type,which has an advantage in environmental and economic
efficiency, generally resists against the sliding of the block
due to the friction of filling geo-materials, it is considered to

be unstable comparing with the retaining wall in wet masonry
type. Thus, it is instructed by adding a projection into eachblock or by using the block with an engagement structure [2]through the method of unifying each block. Thus, it isinstructed to use the method unifying each block by adding a
projection into each block or by using the block with an
engagement structure [2].In addition, an alternative method toenhance the sliding resistance by connecting each block with
furniture has also been studied [3]. However, more effort isstill required to eliminate the anxiety of the stability of blocks.
When enough sliding resistance is achieved in dry masonry
retaining wall compared to wet masonry retaining wall, thesignificant advantages can be expected in relation to
constructive, economic and environmental aspects. Whenachieving an enough sliding resistance in retaining wall withdry masonry which exceeds that in retaining wall with wet
masonry, the significant advantages can be expected inrelation to constructive, economic and environmental aspects.
Corresponding to these concerns, the authors have developed
a reinforced dry masonry block retaining wall in which a newplate is set up at the boundary of upper and lower block.
Corresponding to these concerns, the authors have developed
a dry masonry block retaining wall reinforced the sliding
resistance of each block to set up a new plate at the boundaryof upper and lower block. In this study, a series of verticalstress loading tests were first carried out for investigating theeffect of resistance plate on the sliding stability, and then forevaluating its effectiveness.
2. OVERVIEW OF PROPOSED DRY MASONRYBLOCK STRUCTURE WITH A RESISTANCE
PLATE

611

Fig. 1 A conceptual diagram of a dry masonry block structurewith a resistance plate

Fig. 2 A schematics of the experimental apparatus and model
test specimen

Photo. 1 Earth pressure gage

Photo. 2 Experimental setup

Figure 1 shows a conceptual diagram of a dry masonry block
structure with a resistance plate. Each block used in thepresent structure has a side wall connecting the right and leftof the front wall and rear wall, which is characterized bybottomless plate shown in Fig. 1. Each block used in thepresent structure has a side wall connecting the right and leftof the front wall and rear wall, which is characterized by that
there is no bottom plate shown in Fig. 1. Newly added
resistance plate is perpendicular to the horizontal boundary
surface between the upper and lower block, which is set up by
keeping a spacing from the inner surface of the rear wall.
Then, each inner space of the block and the space behind theblock are densely filled in grain materials. As a result, a
confining layer is produced in the space between theresistance plate and the rear wall, which is named as “rearconfining layer”.
3 OUTLINE OF CONDUCTING MODEL TESTS
Figure 2 shows schematics of the experimental apparatus andmodel test specimen used in this experiment. All of model
tests were conducted by applying the vertical load to the backfilling and then resultantly by giving the horizontal load to the612

Fig. 3 Position of the resistance plate (ex. Case2)
Table 1 Fundamental properties of sand
D 50
0.412mmUc
1.90
Table 2 Experimental conditions
Case
Position of a resistance plate
Case1
No plate
Case2
Back
Case3
Center
Case4
Attached on the back side wall

(a) Case 1 ( No plate)Fig. 4 Relationship between earth pressure and vertical load(b) Case 2 ( Back)(a) Case 1 (No plate)(b) Case 2 (Back)
Fig. 5 Relationship between horizontal displacement and vertical loadback side wall of the block. Scale of a model block made ofacrylic with 15mm in thickness has 220mm front wall width,
200mm depth and 100mm height. Specimen with 300mm
high wall has been created by overlaying a 3-stage block.
Scale of a model block made of acrylic with 15mm inthickness is around 220mm in front wall width, 200mm indepth and 100mm in height. Specimen with 300mm high wall
has been created by overlaying a 3-stage block. In thisexperiment, the steel plate with 2.3mm in thickness, 100mm
in vertical height and 180mm in horizontal width were used
as a model resistance plate.
Horizontal geostatic stress acting on the rear of the specimen
was measured by ultra-compact earth pressure gauge(PDA-200KPA) as shown by Photo 1. Five earth pressure
gages are installed to a plate with 100mm in length, 180mm inwidth and 7mm in thickness, which was bonded to two acrylic
sheets and an aluminum sheet with 1mm thickness. As shownin Photo 1, the pressure gages are vertically attached to thecentral part of the plate at intervals of 22mm. Then, the lateralearth pressure was measured in a total of 15 points byattaching the plate with the pressure gages to the back of eachblock. A typical sea sand is used as filling and backfillingmaterials of the model ground, whose fundamental properties
are summarized in Table 1. In order to prepare the model
ground, all the specimens were prepared by compacting thesea sand under air-dried condition.
Vertical load was applied to the rubber plate put on thesurface of backfill model ground through the water tank inwhich the increasing rate of amount of water is constantly
controlled. In this study, the rubber plate with 220mm width,
200mm in length and 50mm in thickness is set up to 50mm
behind the back side of the block as shown in Photo 2 and theapplicable applied load was limited to be around 4kN.
Also, in order to measure the lateral displacement of eachblock and the vertical displacement on the surface of the top
block, the displacement sensor was set in the center front ofeach block wall. At the same time, two sensors were also put
on the surface of the front and rare sides of the top block.
Figure 3 shows the position of the resistance plate. A series ofexperiments were conducted paying attention to the effect ofthe position on the sliding stability. The position of a
resistance plate selected is summarized in Table 2.
4 RESULTS AND DISCUSSIONS
4.1 Characteristics of mobilized rear earth pressure andhorizontal displacement
Figures 4(a) and (b) show the results of rear earth pressure
acting on the block body without and with resistance plate
against applied vertical loads respectively. Figures 5 (a) and(b) also show the results of the horizontal displacements ofthe block body without and with resistance plate against
applied vertical loads, respectively. Dotted lines in eachfigure indicate the start point of the loading which is around0.59kN. Here, the total force acting on each rear of the block
body (kN) was calculated by multiplying the average value of613

(a) Total rear earth pressure(b) Relative horizontal displacement
Fig. 6 The characteristics of (a) total rear earth pressure and (b) relative horizontal displacement at final loading stage

Fig. 7 Relationship between total rear earth pressure andrelative horizontal displacement at the bottom block

Fig. 8 Relative effect of resistance linked with plate
position, θ defined as a gradient in the relationship between
P H values normalized by γH2B/2 and the relative horizontal
displacement δh normalized by the depth of the block D

the five measured earth pressures by the area of the plate
resistance. It can be confirmed from these figures that thehorizontal displacement and earth pressure at each block haveincreased with the increase of the loading in both cases.
However, paying attention to the increment rate, the rate ofearth pressure in case with resistance plate becomes greatercomparing with that in case without resistance plate. On theother hand, the incremental rate of horizontal displacement incase with resistance plate is kept smaller comparing with thatin case without resistance plate.

considering the results obtained. Figures 6 (a) and (b) aresummarized in the bar diagram to indicate the characteristics
of total rear earth pressure P H and relative horizontal
displacement at final loading stage. The numerical values
shown above the bar diagram is obtained by normalizing themagnitudes of P H in case 2 and 3 by the magnitude in case 3,
which corresponds to the case without any resistance plates. It
is clear that although the magnitudes of P H in the cases withresistance plates become much greater comparing with that incase without any resistance plates, the relative horizontal
displacement becomes smaller. For instance, the relative
horizontal displacement of bottom block in case 3 has been
reduced to less than half. In addition, the effect of resistanceplate on sliding is clearly dependent on the position. Namely,when the resistance plate is set up on the back side of theblock, the effect becomes much greater comparing with thecase whose resistance plate is set up on the middle of theblock.
Figure 7 shows the relationship between relative horizontal
displacement and total rear earth pressure P H at the bottom
block body in three cases. It is confirmed from this figure thatP H values for the case with resistance plate quickly increases
comparing with the case without any resistance plates. A
similar tendency was also confirmed in middle and top blockbodies. Therefore, the incremental rate of P H values against

4.2 Evaluation of effect of resistance plate against sliding
It means that the sliding of the block is suppressed more
effectively when a horizontal displacement of the block body
becomes smaller against a same earth pressure action on thewall. Here, the effect of resistance plate on the sliding of theblock is discussed from the relationship between horizontal
displacement and mobilized earth pressure. In this study, thesum of the earth pressure acting on each block is defined as aforce to slide the bottom surface of the block. For instance,sum of the rear earth pressure action on middle and top blockbodies is used, when focusing on total earth pressure at thebottom of the surface in the middle block body. This isdefined as total rear earth pressure P H . In addition, the relative
horizontal displacement between each block is used for614

relative horizontal displacement is introduced to evaluate theeffect of the resistance plate. Figure 8 shows an index θ which
may indicate the relative effect of resistance linked with plate
position, where θ is defined as a gradient in the relationshipbetween P H values normalized by γH2B/2 and the relative
horizontal displacement δh normalized by the depth of theblock D, where γ is unit weight of soil (= 17.5kN/m3), H isheight of the block specimen and B is width of the block.
Numbers indicated in Figure 8 mean the value which was
calculated by dividing the value of θ for the two other casesby the value in corresponding each block without any
resistance plate. It is indicated that the resistance potential
against sliding at the bottom block in case 3 whose resistanceplate sets up on the back side of the block body becomes
relatively 3.9 times greater. Similarly, the resistance potential
in case 2 whose resistance plate sets up on the middle of theblock body becomes relatively 2.4 times greater. In addition,
the resistance potential against sliding is clearly dependent on
the position of the block body, even if the resistance plate set
up on the same position of the block body. It means that thesliding resistance of the plate is influenced by the induced
overburden pressure.

(a) Top(b) Middle

5 EARTH PRESSURE ACTIING ON THE
RESISTANCE PLATEThe resistance plate has been confirmed in Chapter 4 to beeffective to suppress the sliding of the block body. This isprobably due to earth pressure acting on the resistance plate,which is named as a passive mobilized earth pressure in thisstudy. So, in order to make clear the effect of the position ofresistance plate on the block sliding, the horizontal earthpressures action on the resistance plates were measured incase 2 and case 3, together with case 4 that the resistance plateis attached on the back side of the block.
Figure 9 shows the relationship between the total horizontal
earth pressure acting on the plate and the mobilized horizontal
displacement of the block body, where the total horizontalpressure is estimated by multiplying an average horizontal
earth pressure by an area in the resistance plate. Figures 9(a),(b) and (c) show the results in top, middle and bottom block
body respectively. In Figure 9(c), the mobilized passive total
earth pressure in case 2, 3 and 4 becomes 0.24kN, 0.19kN,
and 0.23kN respectively. On the other hand, the amount ofhorizontal displacement becomes 0.62mm, 0.9mm, and
0.78mm in case 2, 3 and 4 respectively. Based on these results,
it shows that greater the mobilized passive total earth pressure,
relatively smaller the corresponding horizontal displacement.
Conversely, it reflects that smaller the mobilized passive total
earth pressure, relatively greater the corresponding horizontal
displacement. Similar tendency is also indicated in Figures 9(a) and (b). It is mentioned that the effect of a resistance plate
against sliding is due to the mobilized earth pressure action on
the resistance plate. The magnitude of the earth pressure isalso dependent on the overburden pressure by the fillingmaterials in the block. Further considering the results, themobilized earth pressure in case 2 becomes greater than that(c) Bottom
Fig. 9 Relationship between the total rear earth pressure andacting on the plate and mobilized horizontal displacement of
the block bodyof case 4. It is concluded that the resistance plate set up on theback side of the block works well for increasing sliding
resistance of dry masonry retaining structure, comparing withthe case in which the resistance plate was set on the positionattached on a wall at block body. It is considered that this isdue to a constraining effect induced by a granular layer
between the position of the resistance plate and back side ofblock body.
6 CONCLUSIONSIn this study, a series of vertical stress loading tests werecarried out for investigating the effect of resistance plate on
the sliding stability. The main conclusions obtained aresummarized below:
1) Based on the small model test results, it is clearly
confirmed that the resistance plate contributes well to induce
a greater horizontal earth pressure acting on the block body615

and to reduce the horizontal displacement of block body. In
addition, the effect becomes greater when the resistance plate
set up on the back side of the block body, rather than set up on
the middle position of the block.
2) Based on the small model test results, the resistancepotential against sliding at the bottom block in case whose
resistance plate sets up on the back side of the block body
became 3.9 times greater than that of the case without any
resistance plate. Similarly, the resistance potential in casewhose resistance plate sets up on the middle of the block body
became relatively 2.4 times greater.
3) It is confirmed from the characteristics of mobilized earthpressure acting on the resistance plate that the effect of theplate against sliding is due to the mobilized earth pressure
action on it. Furthermore, it is concluded that the resistanceplate set up on the back side of the block works well forincreasing sliding resistance of dry masonry retaining
structure, comparing with the case in which the resistanceplate was set on the position attached on the wall at block
body. One of the reasons is considered that this is due to a
constraining effect induced by a granular layer between theposition of resistance plate and back side of block body.
7 REFERENCES[1] Doro dokou kasetsukozobutsu-ko shishin, Japan roadassociation, 1999, pp.74-75.
[2] Doro dokou kasetsukozobutsu-ko shishin, Japan roadassociation, 1999, p.82.
[3] “Effect of connecting method on dynamic behavior of
concrete block retaining wall”, 29(3), pp. 523-528, 2007

616

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Evaluation and Prediction Method on Neutralization of Supplied Long-term
Hydraulic Concrete Structure
Man-Kwon Choi 1, Yuki Hasegawa2, Shinsuke Matsumoto3, Shushi Sato3 and Tsuguhiro Nonaka4
1
The United Graduate School of Agricultural Science, Ehime University, Japan
2
Graduate School of Integrated Arts and Science, Kochi University, Japan
3
Faculty of Agriculture, Kochi University, Japan
4
Faculty of Life and Environmental Science, Shimane University, Japan
ABSTRACT
Velocity of neutralization is occupied by three factors; W/C ratio in mixture proportion, type of cement and supplied
environmental condition. Hydraulic concrete structures are objective in this study. Influences of supplement condition
to neutralization characteristics were evaluated. As a result, coefficient of neutralization velocity of concrete exposed toair is larger than that of concrete exposed to water. However, as for hydraulic concrete structure, mortar is disappearedand coarse aggregate is exposed from concrete surface in water during long-term supplement. It means that coefficientof neutralization velocity of concrete in water could be an apparent coefficient and actual coefficient was almost samewith that of concrete in air. On the other hand, coefficient of neutralization velocity of concrete canal was smaller thanthat of general engineering structures. Furthermore, neutralization depth and velocity of concrete canal was different
even in same member. Therefore, it seems reasonable to conclude that evaluation of neutralization should be carefulwhen concrete in water is measured.
Keywords: Neutralization, Phenolphthalein method, Neutralization depth, Prediction method, Coefficient of
neutralization velocity

1. INTRODUCTION
Neutralization of concrete generates at anywhere wherecarbon dioxide gas is supplied. This deterioration, commonly
progresses from surface to inner part of concrete, gives
serious damage to reinforced concrete. Investigation anddiagnosis are necessary to carry out from the viewpoint on
restriction of reinforcement corrosion in concrete.
Neutralization depth is the general expression for theevaluation of this deterioration. This depth is indicated withfollowing Eq. (1);y = at b(1)

where y is neutralization depth (mm), t is time (year), a and b
are the constants; especially a is called as coefficient of
neutralization velocity. In general, a constant b is stipulated as
0.5, which was confirmed from many previous studies.
Thus, the progress estimation of neutralization will also be
achieved if there were plural measured data concerned toneutralization depth and supplied year [1].
On the other hand, following Eq. (2) is described for theestimation of neutralization depth when there were nomeasured data;y = (− 3.57 + 9.0 W B ) t(2)

where W/B is the effective ratio of water per binder, W is theunit weight of water and B is the unit weight of effective
binder [1]. However, it is difficult to identify W/B because thedrawings and specifications of most existing old structures
have disappeared during their long-term supplement.
Moreover, progress velocity of neutralization is commonly
occupied by two main factors; W/C ratio in mixture
proportion of concrete (penetration velocity of carbon dioxidegas) and supplied environmental condition of concrete
structures. It means that Eq. (2) is not always available for theprediction of neutralization if neither inspection data nor
drawing and specifications were remained. Furthermore,
Kawanishi et al. have reported that constant a was changed
under some distinctive environmental conditions [2]. As themix, proportion of hardened concrete is impossible toestimate accurately, it is necessary to clarify the influence of
supplemental environment to neutralization of concrete inorder to improve the precision on evaluation and prediction ofneutralization.
Concrete canals for the agricultural use, one of the hydraulicconcrete structures, are supplied long period under severenvironmental condition. At the same time, concrete canals
have one characteristic supplied environmental condition.
The supplied condition of concrete canal is divided into twoparts; the part that sunk into the water (call “water part”) andthe part exposed to the open air (call “air part”) forcomparatively long period. On the whole, it is considered thattransfer velocity of carbon dioxide gas of concrete exposed toliquid phase is quite slower than that of concrete exposed togaseous phase, and the progress of neutralization is possible617

to disregard. However, in case of concrete canal, watercontacted to concrete surface is continually moving, there is apossibility that various ionic compounds become easy todissolve and transfer due to water flow. Furthermore,
exposure of coarse aggregate, one of the distinguishing trait
deterioration of concrete canal, also can affect to theneutralization of concrete.
This research aimed to clarify the influence of different
environmental conditions to the neutralization depth of
concrete experimentally. Objective structures were actual
concrete canals for the agricultural use that were supplied forlong terms. To be specific, the surfaces in water and in air ofconcrete canal were shaved by the rock drill. Phenolphthalein
method was applied and measured the neutralization depth.
Relationship between supplied year and neutralization depth,
as well as the influence of different supplied conditions, in air
and in water, to the progress of neutralization, were
investigated in this study. Finally, applicable prediction
method for the evaluation of neutralization against concrete
canal was examined.
2. RESEARCH SIGNIFICANCE
Neutralization of normal civil engineering structure has beenstudied and their tendency about this deterioration has
clarified sufficiently. In spite, huge number of hydraulicconcrete structures has been constructing in this world, their
specific supplied condition and effect against neutralization
has not focused enough yet. Hydraulic concrete structures are
considered to face the multiple deterioration; not only theneutralization but also the abrasion. It will be necessary toevaluate both deterioration for accurate evaluation andprediction of them. The authors believe that this study will
work for the review on the effect of supplied environment to
the neutralization of concrete.
3. EXPERIMENTAL PROCEDURE
The subject of this study is accumulating the neutralization

data of hydraulic concrete structures in order to grasp thetendency and clarify the influence on supplied environment to
the neutralization. An object of this study was the large
concrete open channel supplied for long periods. Especially aside wall of concrete canal was focused member for theinvestigation because the side wall has both water part and airpart in same member.
3.1 Objective Concrete Canals
Open canal network focused in this study was located in thecenter of Kochi prefecture, Japan. This canal network wasdeveloped from 1968 to 1985 consecutively under prefectural
management for improving the water utilization of paddy
field. In Japan, even though these projects had public
infrastructural aspect, but the part of construction cost shouldbe paid by the beneficiaries (the beneficiary pays principle).
The size of beneficial area was 1,860 hectares paddy fields,
and the length of this main canal network was 36kilometers.
Designed service life of mainstay concrete open channel forthe agricultural use is established for 40 years, and design
criteria strength of concrete was estimated to 21MPa(approximately 200kgf/cm2). Supplied periods ofinvestigated concrete canal were between 23 to 39 years.
Moreover, additional two concrete canals supplied longer
periods, 42 and 47 years, were examined in order to analysis
the relationship between service periods and neutralization
depth in detail.
3.2 Field Investigation Method
Concrete surface was shaved as one of the partial destructive
method for the diagnosis of concrete. Phenolphthalein
method was applied against these shaved phases of concrete.
1.0% concentration of phenolphthalein ethanol solution wasused in this study. Measurement procedure was described asfollows; this solution was sprayed after opening at least
50mm depth of hole at the side wall of concrete canal andblowing off the fine powders of concrete completely. Waiting
for a couple minutes, then, measured the depth of
discoloration boundary from the surface. Phenolphthalein

Fig.2 An Overview on Evaluation of Neutralization
Depth

Fig.1 State of Discolored Condition by
Phenolphthalein Solution
618

ethanol solution is one of a coloring reagent and applies forthe evaluation of neutralization depth of concrete as thissolution change the color from clear into deep red-violet
when the concrete has enough alkalinity. In general,
discoloration generates when the potential of hydrogen (pH)of concrete is higher 8.2 to 10. Authors have also confirmed
the discoloration boundary of this solution and clarified thethreshold value on potential of hydrogen was 10
experimentally [3]. The state of discolored condition by thephenolphthalein solution is shown in Fig.1, and an overview
on evaluation of neutralization depth is shown in Fig.2. The
neutralization depth, distance from surface of concrete todiscoloration boundary is measured using depth gage. Threeholes were drilled at air part and water part in each one
member, and calculated the average depth.
4. RESULTS AND DISCUSSION
4.1 Neutralization Depth and Supplied Periods
Relationship between neutralization depth and supplied
periods are shown in Fig.3. As a whole, neutralization depths
are less than 30mm in spite of their different service periods.
In addition, we cannot see the correlation between them.
Basically the neutralization depth would be larger as the
supplied periods were longer. It is considered that the
progress velocity of neutralization had a great influence to theneutralization depth. Say in other words, it is necessary to pay
attention not only to the service periods but also to thecoefficient of neutralization velocity (call CNV here after) forthe evaluation of neutralization.
Comparing the neutralization depth of air part and water part,
we can confirm big difference between them. There is atendency that the neutralization depth of water part is smaller
than that of air part. Phenomenologically, it would be resulted
that external factors such as the difference on concentration ofcarbon dioxide gas and humidity affect to the neutralization
velocity. However, coarse aggregates were exposed at mostof all water part on the side-wall of concrete canal shown in
Fig.4, and this phenomenon is omnipresent as thedeterioration of hydraulic concrete structures. Neutralization
depth of water part should be evaluated if the equivalence ofevaluation with that of air part was necessary to secure.
Therefore, it is revealed that measured neutralization depth
could not be the enough data for the accurate evaluation of
hydraulic concrete structures if the coarse aggregates were
exposed on their surface.
4.2 Substantial Neutralization Depth of HydraulicConcrete Structures
Described as foregoing paragraph, neutralization depth ofwater part has possibility that its evaluation result was theapparent neutralization depth. According to author’s previous
study concerning to the explosion of coarse aggregate, thisphenomenon would be actualized when 6mm thickness ofmortar was disappeared from the surface of concrete despiteof water cement ratio [4]. It was considered that at least 6mm
thickness of mortar has already abraded if the coarse

Fig.3 Relationship between Neutralization Depth andSupplied Periods
Air Part

Water Part

Fig.4 An Overview of Abraded Surface of ConcreteCanalaggregate were exposed at the side wall of concrete canal.
Relationship between estimated neutralization depth taking
into account the 6mm thickness of disappeared mortar andsupplied periods are shown in Fig.5. Without thisconsideration, 96% of neutralization depth of air part was
higher than that of water part. However, thus, in addition tothis consideration, only 48% of neutralization depth of waterpart could show the same result. As a result, the assumption
concern to the general theory that the neutralization velocity
of water part is smaller than that of air part cannot becomprised all the time against hydraulic concrete structures.
Significant abrasion of concrete canal is observed frequently
at the draft level; the highest frequency water level. Therefore,
evaluation of neutralization at this water level will be also
more important for securing the accuracy on thisdeterioration.

619

Fig.5 Relationship between Neutralization Depth
Considering 6mm Thickness of Disappeared Mortar
and Supplied Periods

Fig.6 Histogram on Coefficient of Neutralization
Velocity of Hydraulic Concrete Structures

4.3 Comparison on Coefficient of Neutralization
Velocity (CNV)
A fundamental equation on the neutralization depth wasshown as Eq. (1). Now CNV was calculated on assumptionthat the neutralization advances in proportion to a square rootof time. In short, CNV was calculated from the measured dataand supplied years. A histogram of CNV was shown in Fig.6,
and their average and standard deviation were shown in Table
Ι. Around 55% of CNV of air part was less than 2.0 in thewhole. However, around 90% of CNV of water part was lessthan 1.0. In addition, if considering the disappeared mortar
thickness in water part was 6mm as an assumption, around
95% of CNV in water part was less than 2.0. Previous study
has reported that the highest frequency range on CNV ofgeneral civil structures (nominal strength: 16 to 21MPa) were
0.0 to less than 2.0. All the results from air part, water part,and water part of additional condition, showed the sametendency with the past study [5]. In the same literature,
percentages of CNV under the range distribution of 0 < CNV< 2, 2 < CNV < 4, 4 < CNV < 6 were around 50%, 35% and15%, respectively. Comparing the same range distribution
between it and our results, CNV of air part of concrete canal
was approximately same or smaller than the result of civilengineering structures; 0 < CNV < 2, 2 < CNV < 4, 4 < CNV< 6 were around 55%, 40% and 5%, respectively.
As for the average and standard deviation results of CNV
shown in Table Ι, average on CNV of air part was three times
larger than that of water part. As well, the standard deviationon CNV of air part was five times larger than that of water
part. In particular, the difference of standard deviation
between air part and water part was generated from the
supplied environmental conditions. It was considered to be
the difference of humidity; the air part was faced to various
conditions of dryness and moisture due to the rainfall andTable Ι Average and Standard Deviation of Coefficient of
Neutralization Velocity
Standard
Average (mm)deviation (mm)Air part

1.92

1.08

Water part

0.60

0.20

Water part(+6mm)

1.65

0.21

solar radiation, on the contrary, water part was more stablecondition rather than air part. These environmental conditions
affected to the difference of standard deviation of them.
4.4 Prediction of Neutralization Considering theDisappeared Mortar
It was considered that the mechanism on progress ofneutralization was different between air part and water part ofconcrete canal. To be more precise, neutralization at the airpart of hydraulic concrete structures advanced with the same
way with general civil engineering structures, however, at thewater part, the exposure of coarse aggregate due to abrasion isgoing on at the same time with the progress of neutralization.
Therefore, it would be impossible to express the equation forprediction of neutralization such as Eq. (1) when the partial
loss of surface is going on such as water part of hydraulicconcrete structures.
For example, as for the predictive equation against sulfate
acid attack, one kind of serious deterioration of concrete,different equations are applied due to their deteriorative
conditions; while the original shape of structure is keeping or
not. In specific terms, deterioration velocity increased inproportion to the square root of time if the shape of original
structure was remaining as it was. On the contrary,
deterioration velocity increased in proportion to the time if

620

the shape was partially loss [6]. Then predictive equation was
assumed to follow to the latter theory, and expressed theequation for the water part could be as follows:y = at(3)

Average and standard deviation of CNV were calculated withthis Eq. (3) and gained 0.29 and 0.05, respectively.
Comparing this result with the result from Eq. (1), CNV was
extremely small. In other words, progress velocity ofneutralization in water part should be smaller than the air part
because the mortar area is becoming smaller due to the
exposure of coarse aggregate, and as a result, penetration
amount of carbonate ion into mortar was decreased relatively.
On the other hand, neutralization depth drawn in Fig.3, theapproximation line should be sloped from right to left.
However, the neutralization depths were constantnevertheless the progressing of time. It was considered thatthe reason was the assumed thickness of disappeared mortar;
constant 6mm thickness of mortar. It meant that the exposureof coarse aggregate was actualized when the 6mm thicknessof mortar was disappeared, but it would be difficult toevaluate the actual disappearance accurately at the presencein case if more than 6mm thickness of mortar had lost. Thus,
it will be possible to evaluate the actual neutralization depth if
the actual disappearance thickness of mortar could beobtained in detail in the future.
5 CONCLUSIONS
Based on the results of this experimental investigation
concerns to neutralization of actual hydraulic concrete
structures, the following conclusions are drawn:1) Comparing the air part and water part of neutralization
depth and coefficient of neutralization velocity, both valuesof water part were smaller than those of air part. Moreover,standard deviation also showed the same tendency with above
result.
2) Those differences between air part and water part were notconfirmed when considering the estimated thickness 6mm ofdisappeared mortar.
3) Coefficient of neutralization velocity of hydraulic concrete
structures was smaller than that of general civil engineering
structures even though the nominal strength was equivalenteach other.

appreciation to Mr. Kenichirou Uchida and Shutaro
Yamasaki, now the staffs of Kochi Prefectural Office, to theirkindly support while they were belonging to master course
student in Kochi University, Japan. A part of this work was
supported by JSPS KAKENHI (22780219) (Grant-in Aid forYoung Scientists (B), A representative: Shushi SATO) andthe authors are also grateful for their financial support to thisresearch work.
8 REFERENCES[1] Japan Society of Civil Engineering, Standard
specifications for concrete structures-2007, maintenance:
JSCE, 2007, pp. 92-73.
[2] Kawanishi T, Hamasaki H and Masuda Y, “Analysis on
the carbonation of concrete based on investigation ofexisting buildings,” J. of Structure and Construction
Engineering, No.608, Oct. 2006, pp. 9-14.
[3] Sato S, Masuma Y, Hasegawa Y, Natsuka I, Aoyama A
and Yokoi K, “Development on specific evaluation
technique for the prediction of neutralization of
concrete,” in Proc. 36th Int. Conf. on OWCS, 2011, pp.
373-380[4] Sato S, Ogata H, Nonaka T and Hattori K, “Fundamental
study on the evaluation of exposed coarse aggregate inconcrete channel,” J. of Japan Concrete Institute, 2008,
pp. 699-704.
[5] Public Works Research Institute, and Nippon Structural
Inspection and Technology Association, Manual for
inspecting soundness level of concrete structures using
nondestructive testing: Gihodo Shuppan Co. LTD., 2003,
pp. 50-55.
[6] Japan Society of Civil Engineering, Standard
specifications for concrete structures-2007, maintenance:
JSCE, 2007, pp. 147-149.

6 FURTHER RESEARCH
It has not clarified the exact progressive process ofneutralization in water part. For establishing the progressive
prediction method, further studies concerns to thedeterioration mechanism on water part of concrete should be
examined, including the relationship between neutralization
and abrasion of hydraulic concrete structures.
7 ACKNOWLEDGMENTS
The authors wish to express their gratitude and sincere621

Second International Conference on Geotechnique, Construction Materials andEnvironment, Kuala Lumpur, Malaysia, Nov. 19-21, 2012, ISBN: 978-4-9905958-1-4

MEASUREMENT OF TENSILE PROPERTIES OF GEOGRIDS
Raid R. Al-Omari1and Mohammed K. Fekheraldin2
23Professor, Civil Engineering Department, Nahrain
University,
Email: tosharaid@yahoo.com, Mobile:
Author1, Author
andBaghdad,
AuthorIraq,
00964-7901699487
Country Najaf, Iraq, Email: Kadum_fekheraldin@yahoo.com,
2
Assistant Lecturer, Civil Engineering Department,Institute,
Kufa University,
Mobile: 00964-7811497218
1

ABSTRACT
Geogrid reinforcement of soil has been successfully used for many years in a wide variety of applications. A huge development
plan is expected in Iraq. The use of modern techniques like geogrid reinforced earth is expected. It is essential to gain
experience in testing methods of geogrids using the available materials produced nationally, in neighboring countries and indeveloped countries. Some manufacturers do not provide sufficient data on the tensile properties of their products.
This paper presents data obtained from a series of laboratory tests performed on various configurations of geogrids. The tests
are conducted to determine the physical properties of various seven types of geogrids, these types of geogrids were
manufactured in Iraq, Iran, China and Britain. The mechanical properties included peak tensile strength, modulus of elasticity,
upper yield strength, lower yield strength, tensile strength, non-proportional extension strength, total extension strength,
fracture elongation, elongation at maximum load and total elongation as well as other physical properties such as aperture size,
mass per unit area, rib thickness, rib width, junction thickness, roll width, mesh type, standard color and polymer type. The
single rib strength test according to ASTM 6637-01 was performed using a computer controlled electronic universal testingmachine. The results were compared with some information available from the manufacturers. Good agreement was foundbetween the results obtained from the present tests and those given by manufacturers.
Keywords: geogrid, tensile strength, modulus of elasticity, elongation, polymer type.

1. INTRODUCTION
Geosynthetics include exclusively man made polymericproducts such as geotextiles, geogrids, geonets,
geomembranes, geosynthetic clay liners, and geocomposites.
Polypropylene,
polyester,polyethylene,
polyamide,
polyvinyl choride, and polystyrene are the major polymers
used to manufacture geosynthetics. It is not the properties ofthe polymers, but the properties of the final polymericproducts that are of interest to civil and environmental
engineering applications. Geosynthetics are used as part of
the geotechnical, transportation, and environmental facilities.
Geosynthetic products perform five main functions:separation, reinforcement, filtration, drainage, andcontainment (hydraulic barrier). However, in mostapplications, geosynthetics typically perform more than one
major function [1].
Geogrids are matrix like materials with large open spaces
called apertures, which are typically 10 to 100mm between
ribs that are called longitudinal and transvers respectively.
The ribs themselves can be manufactured from a number ofdifferent materials, and the rib cross-over joining or
junction-bonding methods can vary. The primary function ofgeogrids is clearly reinforcement [2].
Geogrid, is also defined as a geosynthetic formed by a regular
network of integrally connected elements and possess
apertures greater than 6.35 mm (1⁄4 inch) to allow
interlocking with surrounding soil, rock, earth, and othersurrounding materials to primarily function as reinforcement
[3].

622

Ribs for geogrids are the continuous elements of a geogrid
which are either in the machine (MD) or cross-machine
direction (XMD) as manufactured; Junction is the point
where geogrid ribs are interconnected to provide structuraland dimensional stability.
Geogrid is stiff or flexible polymer grid-like sheets with large
apertures used primarily for stabilization and reinforcementof unstable or weak soil. The general types of geogrid are:1- Unidirectional geogrid.
2- Bidirectional geogrid.
3- Extruded geogrid.
4- Bonded geogrid.
5- Woven geogrid, [2]‐[3].
A significant development plan is expected in Iraq. The use ofgeosynthetics in the construction is anticipated. There is no
local experience in measuring the mechanical properties ofgeogrids, in this work an experience has been built up. Also it
is necessary to establish an idea of the available types ofgeogrids first in Iraq, second in neighboring countries andthird in developed countries in order to be readily aware of theavailable options in the markets.
2. EFFECT OF GEOGRID PROPERTIESThe principal requirements of reinforcing materials arestrength and stability (low tendency to creep), durability, easeof handling, a high coefficient of friction and or adherencewith the soil, together with low cost and ready availability [4].
The selection of fabric for a particular construction
application must necessarily depend upon adequate andsuitable fabric properties and characteristics [5].

Geogrid tests are unique in a number of aspects whencompared with geotextiles. Properties relating to separation,
filtration, drainage, and barrier applications are not included
since geogrids always serve the primary function ofreinforcement [2].
The effective important properties of geogrids may beconcerned in three main properties:
1- Physical properties.
2- Mechanical properties.
3- Endurance properties.
Many of the physical properties of geogrids including theweight (mass), type of structure, rib dimensions, junction
type, aperture size, and thickness can be measured directly
and are relatively straightforward. Other properties that are ofinterest are mass per unit area, which varies over atremendous range from 200 to 1000 g/m2, and percent openarea, which varies from 40 to 95%. The latter suggests that
almost all soils will communicate, or strike-through the planeof geogrids, [2].
The effect of geogrid aperture size on the obtained
enhancement in strength was previously studied. Al-Omari,et al., (1987) investigated the effect of interlocking
mechanism between soil particles and the aperture of themesh using triaxial compression tests. It was concluded thatthe strength enhancement in mesh reinforced sand depends onthe mechanism of interlocking which in turn affected by theratio of aperture size to particle diameter [6].
The mechanical properties of geogrids include the geogrid
stiffness, the peak tensile strength, modulus of elasticity,
upper yield strength, lower yield strength, tensile strength,
non-proportional extension strength, total extension strength,
fracture elongation, elongation at maximum load and total
elongation.
Al-Omari and Hamodi (1991) studied the effect of geogridsof varying stiffness values embedded in clays of different
plasticity indices and they concluded that the reduction in
swell increased with increasing the geogred stiffness [7].
Also Al-Omari, et al., (1995) investigated the effect ofstiffness of geogrids using triaxial test on plastic meshreinforced sand. The sand density, mesh stiffness and thenumber of reinforcing layers were varied and they concluded
that there may be limiting reinforcement stiffness, influenced
by the number of reinforcing layers after which there will be
no further gain in strength [8].
As geogrid are used in critical reinforcement applications,some of which require long service lifetimes, it is generally
necessary to evaluate selected endurance properties [2].
3. Tensile Properties of Geogrids using the Single orMulti-Rib Tensile Method
The single or multi-rib tensile test methods cover thedetermination of the tensile strength properties of geogrids by
subjecting strips of varying width to tensile loading. Threealternative procedures are provided to determine the tensile
strength, as follows:

623

3.1 Method A—Testing a Single Geogrid Rib in Tension(N or lbf).
In this method, a single, representative rib specimen of ageogrid is clamped and placed under a tensile force using a
constant rate of extension testing machine. The tensile force
required to fail (rupture) the specimen is recorded. The
ultimate single rib tensile strength (N or lbf) is thendetermined based on the average of six single rib tensile tests[3].
The initial tendency when assessing a geogrid tensile strength
is to pull a single rib in tension until failure and then to note its
behavior. A secondary tendency is to evaluate the in-isolation
junction strength by pulling a longitudinal rib away from its
transverse rib junction. It is important to state “in-isolation”
since there is no normal stress on the junction; strength test
must be done with the entire geogrid structure contained
within soil embedment, [2].
A single rib tension strength test merely uses a constantrate-of-extension testing machine to pull a single rib to
failure. For unidirectional geogrids, this would most likely bea longitudinal rib. For bidirectional geogrids, both
longitudinal and transverse ribs require evaluation. By
knowing the repeat pattern of the ribs, equivalent wide-width
strength can be calculated. Alternatively, a number of ribs canbe tested simultaneously to obtain a more statisticallyaccurate value for the wide-width strength [3].
An in-isolation junction or node strength test can also be
performed. The test method uses a clamping fixture that grips
the transverse ribs of the geogrid immediately adjacent to andon each side of the longitudinal rib as shown in Fig. 1. The
lower portion of the longitudinal rib is gripped in a standard
clamp, and both clamps are mounted in a tensile testingmachine, where the test specimen is pulled apart. The strengthof the junction, in force units, is obtained [2].
The sample may be taken from the end portion of a rollprovided there is no evidence it is distorted or different fromother portions of the roll. The specimens shall consist of three(3) junctions or 300 mm in length (12 in.), in order toestablish a minimum specimen length in the direction of thetest (either the machine or cross-machine direction). All
specimens should be free of surface defects, etc., not typicalof the laboratory sample. Take no specimens nearer theselvage edge along the geogrid than 1⁄10 the width of thesample. Prepare each finished specimen, as shown in Fig.1, to
contain one rib in the cross-test wide by at least threejunctions (two apertures) long in the direction of the testing,with the length dimension being designated and accurately
cut parallel to the direction for which the tensile strength isbeing measured. The outermost ribs are cut prior to testing toprevent slippage from occurring within the clamps. For thosecases where the outermost ribs are severed, the test results
shall be based on the unit of width associated with the numberof intact ribs [3].

Figure 1: Test fixture measuring in-isolation junction strength
3.2 Method B—Testing Multiple Geogrid Ribs inTension (kN/m or lbf/ft).
A relatively wide specimen is gripped across its entire width
in the clamps of a constant rate of extension type tensile
testing machine operated at a prescribed rate of extension,
applying a uniaxial load to the specimen until the specimen
ruptures. Tensile strength (kN/m or lbf/ft), elongation, andsecant modulus of the test specimen can be calculated frommachine scales, dials, recording charts, or an interfacedcomputer [3].

constructions and reinforced slopes. Geogrids can be divided
into two groups [9]:Stiff geogrids, mostly high density polyethylene (HDPE)
with a monolithic mesh structure.
Flexible geogrids, mostly polyethylene terephthalate (PET)with poly vinyl chloride (PVC) or acrylic coating withmechanically connected longitudinal and transverse
elements.
The main design requirements for the use of geogrids in soil
structures result from the geotechnical design. This includes
the calculation of different failure modes resulting inrequirements for: axial tensile design strength of the geogrid
and maximum strain requirements in the geogrid [9].
In this work the stress-strain properties of various types ofgeogrids (seven types) manufactured in four different
countries are investigated. The following types of geogrids
have been employed in this work: Netlon CE121and Tensar
SS2, manufactured in Britain, a type of geogrid manufactured
in Iraq and a type manufactured in China, SQ12, SQ15 andCE131 manufactured by Pars Mesh Polymer in Iran, Fig. 2
shows the types of geogrids.

Netlon CE121

Tensar SS2

Iraqi geogrid

China geogrid

3.3 Method C—Testing Multiple Layers of Multiple
Geogrid Ribs in Tension (kN/m or lbf/ft)
A relatively wide, multiple layered specimen is gripped
across its entire width in the clamps of a constant rate ofextension type tensile testing machine operated at aprescribed rate of extension, applying a uniaxial load to thespecimen until the specimen ruptures. Tensile strength (kN/mor lbf/ft), elongation and secant modulus of the test specimen
can be calculated from machine scales, dials recording charts,
or an interfaced computer [3].
4. The used geogrids
With BS 8006, FHWA-SA-96-071, and independentcertification of product characteristics by the British Board of
Agreement, a sound base for the use of reinforcing products,
whether flexible or stiff, is now available. A range of geogrid
products with varying strength and made in different ways is
now available from which users can select the product best
suited to their purpose. The strength of the geogrids varies
between 20 and 250 kN/m, and they are used in both road

624

PMP SQ12PMP SQ15

PMP CE131

Figure 2: The various types of geogrids used in the work
5. Experimental Results
In the present work a large number of samples (90 samples)
were tested by computer controlled electronic universal
testing machine and this machine give the information as
follow; peak tensile strength, modulus of elasticity, upper
yield strength, lower yield strength, tensile strength,
non-proportional extension strength, total extension strength,
fracture elongation, elongation at maximum load and total
elongation after completing test for each geogrid type. Figure
3 shows sample test using the computer controlled electronic
universal testing machine and Fig. 4 demonstrates theinformation given from this testing machine.

Table 1: The physical and mechanical properties of Netlon
CE121 geogrid
The physical properties
Property
Data
Mesh type
DiamondStandard color
Black
Polymer type
HDPE
Packaging
Rolls
Dimensional Properties
Property
Unit
Data
Aperture size(MD/XMD)

Mass per unit area
g/m2
Rib thicknessmm
Rib widthmmJunction thicknessmmRoll widthmRoll lengthm
The Mechanical Properties
Peak Tensile StrengthkN/mElastic modules
GPa
Upper yield strength
MPa
Lower yield strengthMPa
Tensile strength
MPa
Non-proportional extension
MPastrength
Total extension strength
MPa
Fracture elongation%
Elongation at maximum load%
Tota elongation
%

6*8
740
1.6/1.45
2/2.75
2.75
4
50
6.4
0.39
5
5
9
6
5
-99
6
11

Load (kN)

Figure 3: Computer controlled electronic universal testingmachinemm

Figure 4: Information given after test from computer
controlled electronic universal testing machine
The properties of the geogrids employed in this work are as
follow;
Netlon CE 121
This geogrid was manufactured by the British Company
Netlon ltd. The physical and the presently determinedmechanical properties of this type of geogrid are summarizedin Table 1. The results of single rib test, as determined in thepresent work, are shown in Figs. 5a and b.

625

Deformation (mm)Figure 5a: The load-deformation relationship of
CE121 Netlon geogid

Load (kN)Stress (MPa)Deformation (mm)Figure 6a: The load-deformation relationship of Tensar
SS2 geogrid in XMD direction

StrainFigure 5b: The stress-strain relationship of CE121 Netlon
geogridThis geogrid was manufactured by the British Company
Netlon ltd. The physical and the presently determinedmechanical properties of this type of geogrid are summarizedin Table 2. The results of single rib test, as determined in thepresent work, are shown in Figs. 6a, b, c and d.
Table 2: The physical and mechanical properties of Tensar
SS2 geogrids
The physical properties
Property
Mesh type
Standard color
Polymer typePackaging

Stress (MPa)

Tensar SS2

Data
square
Black
HDPE
Rolls

StrainFigure 6 b: The stress-strain relationship of Tensar SS2
geogrid in XMD direction

Unit

Data

Aperture size(MD/XMD)
mm
Mass per unit area
kg/m2
Rib thickness MD/XMD
mm
Junction thicknessmmLongitudinal rib width lw
mmTransverse rib width twmmRoll widthmRoll lengthm
The Mechanical Properties
Peak Tensile Strength MD/XMD
kN/mElastic modules MD/XMD
GPaUpper yield strength MD/XMD
MPa
Lower yield strength MD/XMD
MPa
Tensile strength MD/XMD
MPaFracture percentage elongation MD/XMD%
Percentage elongation at maximum load%
MD/XMD
Total percentage elongation MD/XMD
%

28/40
0.3
1.2/1.1
3.9
3
3
4
50
14.4/28.2
0.57/0.99
1 /3
1/3
24/30.7-98/-98
3.5/2.9

Load (kN)

Dimensional Properties
Property

5/4.25

Deformation (mm)Figure 6c: The load-deformation relationship of Tensar SS2
geogrid in MD direction

626

StrainFigure 6d: The stress-strain relationship of Tensar SS2 geogrid
in MD direction
The results of the present tests reasonably agree with results
provided by the manufacturer, Tensar specification of geogrid
SS2 in mechanical properties of single rib test. The results
provided by the manufacturer are17.5 kN/m and 31.5 kN/mfor MD and XMD respectively, while the values from thesingle rib test of Tensar SS2 by computer controlled
electronic universal testing machine are 14.4 and 28.2 kN/M
for MD and XMD respectively. The percentages ofagreement are 82.3% and 89.5 for MD and XMD
respectively.

GPa
MPa
MPa
MPa
%
%
%

0.083
0
0
7.5
-99
9
21.1

Load (kN)Stress (MPa)Elastic modules
Upper yield strength
Lower yield strength
Tensile strength
Fracture percentage elongation
Percentage elongation at maximum load
Total percentage elongation

Deformation (mm)Figure 7a: The load-deformation relationship of Iraqi geogrid

A geogrid product manufactured in Iraq is employed. The
physical and presently determined mechanical properties areillustrated in Table 3. Figure 7a and b demonstrates theobtained stress-strain and load-deformation relationships.
Table 3: The physical and mechanical properties of Iraqi
geogrid

Stress (MPa)The Iraqi Geogrid

The physical properties
Property
Mesh type
Standard color
Polymer typePackaging

Data
Diamond
Blue
HDPE
Rolls

StrainFigure 7b: The stress-strain relationship of Iraqi geogrid

Dimensional Properties
Property

Unit

Data

mm
g/m2
mm
mm
mm
m
m
The Mechanical Properties
Peak Tensile Strength
kN/m

21*18
255
1.4/2.25
3.3
1.6/2
1.2
30

Aperture size
Mass per unit area
Rib thickness
Junction thickness
Rib width
Roll width
Roll length

1.6

627

The China Geogrid
A geogrid produced in China is employed. The physicaland the presently determined mechanical properties areillustrated in Table 4. Figure 8a and b demonstrates theobtainedstress-strain
andload-deformation
relationships.

Table 4: The physical and mechanical properties of China
geogrid

Property
Aperture size
Mass per unit area
Rib thickness
Junction thickness
Rib width
Roll width
Roll length

Data
Diamond
Green
HDPE
Rolls
Dimensional Properties
Unit

mm
g/m2
mm
mm
mm
m
m
The Mechanical Properties
Peak Tensile StrengthkN/mElastic modules
GPa
Upper yield strength
MPa
Lower yield strengthMPa
Tensile strength
MPa
Fracture percentage elongation%
Percentage elongation at maximum%
load
Total percentage elongation
%

Data
12*15
283
1.2/1.1
1.9
1.4/1.25
1.2
30
1.1
0.03
------------6.5
-99.5
22.7
31.9

Stress (MPa)The physical properties
Property
Mesh type
Standard color
Polymer typePackaging

StrainFigure 8b: The stress-strain relationship of China geogrid
PMP SQ12
This geogrid is manufactured by the Iranian company
Pars Mesh Polymer. The physical and the presently
determined mechanical properties of this geogrid aresummarized in the Table 5. The results of single rib test,
as determined in the present research are shown in Figs.
9a and b.
Table 5: The physical and mechanical properties of SQ12
geogrids
The physical properties

Load (kN)

Property
Mesh type
Standard color
Polymer typePackaging
Property

Deformation (mm)Figure 8a: The load-deformation relationship of China
geogrid

628

Data
Square
Green
HDPE
Rolls
Dimensional Properties
Unit

Aperture size
mm
Mass per unit area
g/m2
Rib thicknessmmJunction thicknessmmLongitudinal rib width lw
mmTransverse rib width twmmRoll widthmRoll lengthm
The Mechanical Properties
Peak Tensile StrengthkN/mElastic modules
GPa
Upper yield strength
MPa
Lower yield strengthMPa
Tensile strength
MPa
Fracture percentage elongation%
Percentage elongation at%
maximum load
Total percentage elongation
%

Data
12*12
318
1.7
1.6
1.3
1.6
1.2
30
0.25
0.28
--------------2
-99.5
1
5.5

Unit

Data

Aperture size
mm
Mass per unit area
g/m2
Rib thicknessmmJunction thicknessmmLongitudinal rib width lw
mmTransverse rib width twmmRoll widthmRoll lengthm
The Mechanical Properties
Peak Tensile StrengthkN/mElastic modules
GPa
Upper yield strength
MPa
Lower yield strengthMPa
Tensile strength
MPa
Fracture percentage elongation%
Percentage elongation at maximum load%
Total percentage elongation
%

24*24
385
2
3.3
2.5
2.2
1.2
20

Load (kN)

Property

Stress (MPa)

Load (kN)

Deformation (mm)Figure 9a: The load-deformation relationship of SQ12
geogrid2.41
0.١2/0.54----/-------/---2
-9٧
0.٥
2

Deformation (mm)Figure 10a: The load-deformation relationship of SQ15
geogrid

StrainFigure 9b: The stress-strain relationship of SQ12 geogridThis geogrid is manufactured by the Iranian company Pars
Mesh Polymer. The physical and the presently determinedmechanical properties of this geogrid are summarized inTable 6. The results of single rib, as determined in the present
tests are shown in Figs. 10a and b.
Table 6: The physical and mechanical properties of SQ15
geogrids

Stress (Mpa)

PMP SQ15

The physical properties
Property
Mesh type
Standard color
Polymer typePackaging

Data
Square
Green
HDPE
Rolls

StrainFigure 10b: The stress-strain relationship of SQ15 geogrid

Dimensional Properties

629

PMP CE131
This geogrid is manufactured by the Iranian company Pars
Mesh Polymer. The physical and the presently determined
Table 7. The results of single rib test, as determined in thepresent research are shown in Figs. 11a and b.
Table 7: The physical and mechanical properties of CE131
geogrids

Stress (MPa)mechanical properties of this geogrid are summarized in

The physical properties
Property
Mesh type
Standard color
Polymer typePackaging

Data
Square
Green
HDPE
Rolls

StrainFigure 11b: The stress-strain relationship of CE131 geogrid

Dimensional Properties
Property

Unit

Data

Aperture size
mm
Mass per unit area
g/m2
Rib thicknessmmJunction thicknessmmLongitudinal rib width lw
mmTransverse rib width twmmRoll widthmRoll lengthm
The Mechanical Properties
Peak Tensile StrengthkN/mElastic modules
GPa
Upper yield strength
MPa
Lower yield strengthMPa
Tensile strength
MPa
Fracture percentage elongation%
Percentage elongation at maximum load%
Total percentage elongation
%

34*34
429
2.8
5
3.1
3.1
1.2
30
1.92
0.3٢
--------------1
-9٨.0
1.5
8.5

6. Discussion of test results
This study was carried out to investigate the parameters thatgovern the behavior of geogrids. The most significant results
obtained from the present work are summarized in Table 8. It
is obvious that Tensar SS2 and Netlon CE121 possess thehigher tensile strength and modulus of elasticity. All othergeogrids may be used for temporary structures or wheremodest strength is accepted.
Table 8: Results of various geogrids
Geogrid

Peak

Modulus of

Rib

Mass

Rib

type

tensileelasticity

Thickness

per

width

strength

MD/XMD

MD/XMD

unit

Lw/tw

MD/XMD

(GPa)

(mm)

area

(mm)

2

(g/m )(kN/m)
Netlon6.4

0.39

1.6/1.45740

2/2.7514.4/28.2

0.57/0.99

1.2/1.1

300

3/3

1.6

0.083

1.4/2.25

255

1.6/2

1.1

0.03

1.2/1.1

283

1.4/1.25

0.25

0.28

1.7

318

1.3/1.6

2.41

0.12/0.542385

2.5/2.2

1.92

0.32

2.8

429

3.1/3.1

CE121
TensarSS2

Load (kN)

Iraqi
geogridChinageogrid
PMP
SQ12
PMP
SQ15

Deformation (mm)Figure 11a: The load-deformation relationship of CE131
geogrid

630

CE131

7. CONCLUSION
Based on this experimental study the following conclusions
may be drawn:
1- The geogrids Netlon CE121 and Tensar SS2 havetensile strength and elastic modulus higher thanother geogrids made by different manufactures.
2- The tensile strength and elastic modulus are notdependent on the density of materials but dependenton type of polymer used and method of manufacture.
3- The dimensional properties such as rib thickness,junction thickness, longitudinal and transverse rib
width of geogrid play important role in themechanical properties such as tensile and elastic
modulus.
4- The effect of tensile strength and stiffness is moresignificant than elastic modulus when geogrids areused as reinforcement in the soil.
.
8. ACKNOWLEDGMENTS
The experimental part of the work was conducted in thelaboratory of Materials Engineering Department in College ofEngineering of Kufa University. Special thanks are presented
to Dr. Ali Sabaa Hummoud, the chairman of Materials
Engineering Department and deepest wishes to Miss. Batool
Hatim, responsible on laboratory of materials engineering.
9. REFERENCES[1] Hoe I. Ling, “Civil and Environmental Applicationsof Geosynthetics,” Reinforced
Soil
Engineering.
Advances in Research and Practice, 4th ed. vol. 1, Hoe I.
Ling, Dov Leshchinsky and Fumio Tatsuoka, Ed. Libraryof Congress Cataloging: Publisher’s, 2003, pp. 14–31.
[2] Roberrt M. Koerner, Designing with Geosynthetics,
Library of Congress Cataloging: Publisher’s, 2005, Ch. 3.
[3] ASTM D6637-01-2009, “Standard Test Method for
Determining Tensile Properties of Geogrids by the Single
or Multi-Rib Tensile Method”.
[4] Colin JFP Jones, Earth Reinforcement and Soil Structure,
British Library Cataloguing: Publisher’s, 1985, Ch. 5.
[5] Roberrt M. Koerner and Joseph P. Welsh, Construction
and Geotechnical Engineering Using Synthetic Fabrics,
Library of Congress Cataloging: Publisher’s, 1980, Ch. 4.
[6] Al-Omari, R.R, AL-Dobaissi, H.H. and Al-Wadood,
B.A.”Inextensible geomesh included in sand and clay”.
Proceedings, International Symposium on Prediction and
Performance in Geotechnical Engineering, Calgary, 1987,
pp. 155-160.
[7] Al-Omari, R.R. & Hamodi F. J. “Swelling ResistantGeogrid- A new Approach for the Treatment of Expansive
Soils” J. of Geotextiles and Geomembranes, 1991, , pp.
295-317.
[8] Al-Omari, R.R., Nazhat, Y.N. & AL-Dobaissi, H.H.
“Effect of Stiffness and Amount of Reinforcement on
Strength of Sand” J. of Engineering Geology, 1995, pp.
363-367.

631[9] Voskamp W.” Performance Properties of Geogrids”
Reinforced Soil Engineering. Advances in Research andPractice, 4th ed. vol. 1, Hoe I. Ling, Dov Leshchinsky and
Fumio Tatsuoka, Ed. Library of Congress Cataloging:
Publisher’s, 2003, pp. 32-48.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Modeling the Swelling strain and pressure of Weak Rockusing Adaptive Network-Based Fuzzy Inference System
Ramin Doostmohammadi
Mining Engineering Department, University of Zanjan, Iran
ABSTRACT
Some of civil engineering projects (such as dam construction, excavation of road way tunnels) or mining projects (suchas underground excavation of coal and gypsum ores) are performed in vicinity of weak rocks with swelling potential.
Therefore, the swelling strain and swelling pressure of such rocks can develop under different boundary conditions,
simultaneously. This paper uses the information of swollen clay content and different boundary stiffness in combination
with Adaptive Network-Based Fuzzy Inference System (ANFIS) to predict the swelling pressure and strain of weak
rock. The result of this method is compared with the previously presented method, Artificial Neural Network (ANN).
Differing from ANN, ANFIS is transparent rather than a black box.
Keywords: Swelling strain, swelling pressure, Adaptive Network-Based Fuzzy Inference System

In an excavation without any support system, swelling rockmoves inside and causes problems associated with reduced
cross sectional area. When excavation is reinforced by
support systems, swelling will usually cause considerable
damages to the support system. The pressure from swelling
rocks to the support system and lining of undergroundopenings has been the focus in recent years.
Basma et al. (2000) implemented ANNs for modeling thetime dependent swelling of expansive soils [1]. They
attempted to implement sequential ANN for modeling thetime dependent swelling strain of expansive soils. Soilsamples with varying properties were selected and tested for
expansion under three different initial applied pressures (25,100, and 200 kPa) to develop the database used for training
and testing the neural network. The input parameters used in
the network included the soil initial dry unit weight, watercontent, initial applied pressure, percent clay content,
plasticity index and the percent swell at time i. Different
models based on the ANN approach were developed for
determination of the swelling parameters of some clayey soils
present in the Oltrepo Pavese area by Meisina and Najjar(2004). Some models were developed to estimate the swellingpressure and others were employed to estimate the swellingstrain. A different number of variables (atterberg limit, clay
content, methylene blue absorption, water content, dry
density and soil suction) were selected to represent the inputs
for the models [2]. Moosavi et al. (2006) modeled the cyclic
swelling behavior of mudstone using ANNs. Each sample has
its own swell model and a general model for all studied
samples is lacking. The duration of a swelling cycle is divided
to twelve equal points. The swell pressure of a desired point ispredictedaccording to its four previous points [3]. In a recent
research, Doostmohammadi et al. (2008) have used ANN for
predicting the swelling strain and pressure of mudstone under
different boundry conditions [4]. The input data for theirmodel are the result of some standard laboratory tests thathave been proposed by ISRM and Madsen [5]-[6]-[7].
2. SWELLING TESTS AND SAMPLE
CHARACTERIZATION
Mudstone cores for experimental study were obtained from
drilled holes at Masjed-Soleiman Underground hydroelectric
power plant (UHEPP). Table 1 shows the mineralcomposition of six samples taken from mudstone at different
locations of Power House Cavern (PHC).
2.1 Free swelling strain test
This test is intended to measure the swelling strain developed
when an unconfined, undisturbed rock specimen is immersed
in water [7]. Such conditions happen in ground surface andshallow trenches. Some results of this test on mudstone
samples are shown in Fig.1.

632

9

Sa.3

8

Sa.6
Sa.5

7

Swelling strain (%)

1. INTRODUCTION

Sa.4
Sa.2

6
5
4
3
2
1
0
0

1

2

3

4

5

6

7

8

Time (Day)Fig.1. The result of the free swelling tests for mudstone
samplesTable 1. The mineralogy of mudstone samples.
Sample number
Montmorillonite (%)
Illite (%)
Kaolinite (%)
Chlorite (%)
Quartz (%)
Feldespar (%)
Calcite (%)
Dolomite (%)

Sa.1
5.9
4.2
7.9
8.8
11.3
10.2
36.9
14.8

Sa.2
3.6
8.2
3.6
6.4
10.3
12.5
39.8
15.6

Sa.3
3
8.5
0
4.5
14.2
6
48.7
15.1

2.2 Semi confined swelling strain test
This test is intended to measure the axial swelling strain undera constant axial load of a radially confined, undisturbed rockspecimen when immersed in water [7]. Examples of suchconditions are tunnel inverts or swelling ground underfoundations. Some results of this test on mudstone samples
are shown in Fig. 2.
9

Sa.5
3.8
10.3
3.4
7
12.1
3.4
49.6
10.4

Sa.6
2.5
7.6
4.1
2.7
18
3.1
49.7
12.3

2.4 Semi confined swelling strain and pressure test
This test is intended to measure the axial swelling strain andpressure of specimen in contact of deformable axial boundary
condition. An example of such conditions is the roof oftunnels supported by deformable reinforcements. Some
results of this test on mudstone samples are shown in Fig.4
and Fig.5.

Sa.6 (under 20.4 kPa)

8

Sa.3 (under 5.4 kPa)

0.6

Sa.5 (under 5.4 kPa)

7

Sa.2 (Stiffness=3850 kPa/mm)

Sa.4 (under 5.4 kPa)

6

Sa.2 (under 5.4 kPa)

5

Sa.4 (under 0.12 kPa)

Sa.1(Stiffness=17000 kPa/mm)

0.5Sa.5 (under 0.12 kPa)
Sa.1 (under 100 kPa)

4

Sa.1 (under 199 kPa)3

Sa.1 (under 0.12 kPa)

2
1

Sa.3 (Stiffness=3660 kPa/mm)
Sa.4 (Stiffness=7785 kPa/mm)

Swelling strain (%)

Swelling strain (%)

Sa.4
2.8
5.5
2.8
5.6
15.25
3.4
49.6
15.05

0.4
0.3
0.2

0
0

1

2

3

4

5

6

7

8

0.1

Time (Day)Fig.2. The result of the semi confined swelling strain test for
mudstone samples

00

Swelling pressure (kPa)

Sa.1
Sa.3

700

3

4

5

6

7

8

Fig.4. The result of the variation of the swelling strain underconstant axial stiffness

500

Swelling pressure (kPa

Sa.2

800

2

Time (Day)

2.3 Swelling pressure test
This test is intended to measure the pressure necessary toconstrain an undisturbed rock specimen at constant volume
when it is immersed in water [7]. Such conditions happen in
lined tunnels containing swelling rock beds. Some results ofthis test on mudstone samples are shown in Fig. 3. The highswelling pressure is reffered to osmotic or double layerswelling as reported by seedsman [8].

9001Sa.2 (Stiffness=3850 kPa/mm)

450Sa.1 (Stiffness=17000 kPa/mm)

400

Sa.3 (Stiffness=3660 kPa/mm)
Sa.4 (Stiffness=7785 kPa/mm)

350
300
250
200
150
100

Sa.4

600

50

500

0

400

0

300

1

2

3

4

5

6

7

8

Time (Day)200

Fig.5. The result of the variation of the swelling pressure
under constant axial stiffness

100
0
0

1

2

3

4

5

6

7

8

Time (Day)Fig.3. The result of the swelling pressure test633

3. DEVELOPMENT OF ANFIS FOR MODELING THE
SWELLING POTENTIAL
3.1 Concept of ANFISThe fuzzy logic approach is based on the linguistic uncertain
expression rather than numerical uncertainty. The main
problem with this approach is that there is no systematic
procedure for a design of fuzzy controller. Basically a fuzzy
inference system (FIS) is composed of five functional blocks(Fig. 6).
- A rule base containing a number of fuzzy if-then rules;- A database which defines the membership functions of the
fuzzy sets used in the fuzzy rules;- A decision-making unit which performs the inference
operation on the rules;

Where x (or y) is the input to the node; Ai (or B i-2 ) is a fuzzy
set associated with this node, characterized by the shape ofthe MFs in this node and can be any appropriate functions thatare continuous and piecewise differentiable such as Gaussian,
generalized bell, trapezoidal and triangular shaped functions.
Layer 2: Every node in this layer multiplies the incoming
signals.
Layer 3: The ith node of this layer, labeled as N, computes thenormalized firing strengths.
Layer 4: Node i in this layer computes the contribution of the
ith rule towards the model output, with the following nodefunction:

OPi 4 = wi f i = wi ( pi x + qi y + ri )(5)where wi is the output of layer 3 and {pi, qi, ri} is theparameter set.
Layer 5: The single node in this layer computes the overall
output of the ANFIS as:∑w f
OP = Overall _ output = ∑ w f =
∑w
i

5

i

i

i

i

i

i

(6)

i

i

Fig.6. The general structure of the fuzzy Inference System- A defuzzification interface which transform the fuzzy
results of the inference into a crisp output.
Jang (1993) introduced a novel architecture and learning
procedure for the FIS that uses a neural network learningalgorithm for constructing a set of fuzzy if–then rules withappropriate membership functions from the stipulated
input–output pairs. This procedure of developing a FIS using
the framework of adaptive neural networks is called anadaptive network-based fuzzy inference system [9].
ANFIS is a Sugeno-type FIS. The general structure of the
ANFIS is presented in Fig.7. It is assumed that the FIS has
two inputs x and y and one output z. suppose that the rule basecontains two fuzzy if-then rules of Takagi and sugeno’s type:
If x is A 1 and y is B 1 Then f1 =p 1 .x+q 1 .y+r 1(1)
If x is A 2 and y is B 2 Then f2 =p 2 .x+q 2 .y+r 2(2)
Where A1, A2 and B1, B2 are the membership functions for
inputs x and y, respectively; p 1 , q 1 , r 1 and p 2 , q 2 , r 2 are theparameters of the output function. Fig.7(a) illustrates the
fuzzy reasoning mechanism for this Sugeno model to derive
an output function (f) from a given input vector [x,y]. The
corresponding equivalent ANFIS architecture is presented in
Fig.7(b), where nodes of the same layer have similarfunctions. The functioning of the ANFIS is as follows:
Layer 1: Each node in this layer generates membership grades1of an input variable. The node output OPi is defined by:

OPi = µ Ai ( x) for i=1, 2(3)

OPi = µ B (i − 2 ) ( x) for i=3, 4

(4)

1

1

Fig.7. Schematic of fuzzy and neurofuzzy paradigm: (a)fuzzy
inference system and (b)equivalent ANFIS architecture
The parameters for optimization in an ANFIS are the premiseparameters {a i , b i , c i }, which describe the shape of the MFs,
and the consequent parameters {p i , q i , r i }, which describe theoverall output of the system. The basic learning rule of an
adaptive network, the back propagation algorithm which isbased on the gradient descent rule, can be successfully
applied to estimate these parameters[10].
3.2 ANFIS model development
The swell potential tests according to part 3 were performed
to develop and assess the ANFIS model. Swelling potential
tests including 354 data points are used for training (283 data

634

Table 2. Maximum swelling strain and pressure using ANFIS and ANN models
Maximum swelling strain(%)

Maximum swelling pressure(kPa)

Real Data

ANN

ANFIS

Real Data

ANN

ANFIS

Test 1

0.45

0.49

2.06

382

448

546

Test2

0.001

0.001

0.00

737

730

607

Table 3. Correlation coefficient and error in predication of swelling strain and pressure
Time dependent swelling strain(%)

Time dependent swelling pressure(kPa)

Error

Error

Correlation coefficient

Correlation coefficient

ANN

ANFIS

ANN

ANFIS

ANN

ANFIS

ANN

ANFIS

Test 1

0.102

3.57

0.963

0.845

0.174

0.43

0.984

0.960

Test 2

0.00

1.00

0.998

0.998

0.0095

0.176

0.978

0.954

points) and testing (71 data points) the model. The training
and testing data points include 80% and 20% of the total data
points are selected randomly from the total data points.
The input parameters are selected according to the important
affecting factors in swelling potential and boundary
conditions. Five parameters are considered for modeling theswelling potential: time at which swell potential is measured,
stiffness in radial direction, percentage of swollen clays(montmorillonite) content, initial dead pressure in axial
direction and stiffness in axial direction. The output
parameters are swelling strain and swelling pressure.
In the present study, two triangular membership functions
have been assigned to each input variable. Each input variable
is classified into two fuzzy categories with linguistic
attributes. The initial values of the premise parameters are set
in such a way that the centers of the MF are equally spaced
along the range of each input variable.

The model structure is implemented using the fuzzy logic
toolbox of MATLAB software package.
The hybrid algorithm used in the present study for optimizing
the parameters allows a fast identification of parameters andsubstantially reduces the time needed to reach convergence.
Fig.8 shows the final MFs for ANFIS pressure model after
training.
The ANFIS model is tested using a matrix of 71 samples
obtained from different laboratory testing groups. These data
points were not used in training the model. The ANFIS modelis used to predict the swelling strain and pressure. The realand predicted maximum strain and pressure for two test
dataset are presented in Table 2.
The performance of the ANFIS model is evaluated bystatistical indexes and is compared to that of ANN method.
The error and correlation coefficient are determined asfollows:

Error = (1

1 N
∑ ( xi − yi )2 ) 2
N i =1(7)

N

Corr =∑ ( x − x).( y − y)
i =1

i

i

N

N

i =1

i =1

(∑ ( xi − x) 2 .∑ ( y i − y ) 2 )

1
2

(8)

Where x=forecasted values, y=measured values, x =mean of
Fig.8. Final MFs for swelling pressure after training

forecasted values, y = mean of measured values and N=
number of data points.

635

The correlation coefficient and error for predicted strain andpressure using ANFIS and ANN model are presented in Table
3, also.
It is observed that ANFIS predicts the swelling potential withhigher error and lower correlation coefficient in comparison
with ANN.
4 CONCLUSIONSa) The results of standard swelling potential tests arecorrelated using ANFIS method. The ANFIS uses the time atwhich swell potential is measured, stiffness in radial
direction, percent of swollen clay content, initial deadpressure in axial direction, stiffness in axial direction as input
parameters to develop the model. It was shown that ANFIS
model can be employed to simulate the swelling potential of
mudstone.
b) The proposed model can be used to determine the swelling
response of rocks under different boundary conditions (freeradial boundary, rigid radial boundary, free axial boundary,rigid axial boundary and deformable axial boundary). The
result of this model can be applied to predict the interaction ofswelling rocks and support systems in undergroundexcavations which usually have argillaceous boundaries.
c) The difference in the maximum swelling strain obtained
from the two methods is large as given in table 2. It has been
observed that ANN (which has been found in literature)outperforms ANFIS. Differing from ANN, ANFIS istransparent rather than a black box.

[8] Seedsman RW, “Comprehensive rock engineering.
characterizing clay shales,” Vol. 3, John A. Hudson:
Pergamon Press, 1993, chap.7, pp. 151-164.
[9] Jang JSR, “ANFIS: Adaptive network based fuzzy
inference system,” IEEE Trans Syst Man Cybernet, Vol.23,
1993, pp.665–83.
[10] Jang, JSR, Sun, CT, “Neurofuzzy modeling and control,”
Proceeding of IEEE, 1995, pp. 378–406.

4. REFERENCES[1] Basma AA, Barakat SA, Omar M, “Modeling timedependent swell of clays using sequential artificial neuralnetworks,” Environmental & Engineering Geoscience, Vol.9,
2003, pp.279-288.
[2] Meisina C, Najjar Y, “Assessing the swelling /shrinkage
potential of Italian soils via neural network approach, ”
Proceeding of 9th Symposium on Numerical Models inGeomechanics, 2004, pp.187-193.
[3] Moosavi M, Yazdanpanah MJ, Doostmohammadi R,“Modeling the cyclic swelling pressure of mudrock using
artificial neural network,” Engineering Geology, Vol.87,
2006, pp.178-194.
[4] Doostmohammadi R, Moosavi M, Araabi BN,“Modelling time dependent swelling potential of mudrockusing laboratory tests and artificial neural network,” Mining
Technology, Vol.117, 2008, pp.32-41.
[5] ISRM, “Characterisation of Swelling Rock,” Pergamon
Press, 1983, pp. 49-75.
[6] ISRM, “Suggested methods for laboratory testing ofargillaceous swelling rocks,” International Journal of RockMechanics and Mining Sciences and Geomechanics
Abstracts, Vol. 26, 1989, pp.415-426.
[7] Madsen FT, ”Suggested methods for laboratory testing ofswelling rocks”, International Journal of Rock Mechanics andMining Sciences and Geomechanics Abstracts,” Vol.36,
1999, pp.211-225.
636

Centrifuge Modeling Of Reinforced Embankments On Soft Foundation
Ali Sobhanmanesh1, Nurly Gofar2
University Teknologi Malaysia, Malaysia

ABSTRACTA series of centrifuge tests were fulfilled at small centrifuge equipment of national university of Malaysia (UKM) on
reinforced and unreinforced embankments built over soft and stiff foundations to observe the effect of reinforcement,
investigate the reinforcement mechanism and deformation behavior of models. Five different cases were analyzed
based on different strength of foundation and embankment soil. A proper textile used as a reinforcement material andclayey-sand used as fill material. Kaolin and dry sand used as soft and stiff foundations respectively. The results showthat surface settlement and deformation behavior of reinforced models are changed and reduced significantly compare
to unreinforced models.
Keywords: Reinforced Embankment, Centrifuge test, Soft foundation
1

1. INTRODUCTION

2.

Many embankments constructed on soft grounds are
susceptible to failure and large settlements [1]. Thesestability problems of embankments on soft soils occur due
to high compressibility and low shear strength of soil. Manyof the ground improvement techniques, which have beenused in the past to increase the shear strength of the soil, aretime-consuming and uneconomical. Hence an alternative
method such as soil reinforcing by geosynthetics is needed
to solve this problem. The emergence of geosynthetic
reinforcement in recent times has revolutionized theconcept of ground improvement. The idea behind thistechnique is to make use of the tensile strength of thereinforcement to limit the spreading of the embankment andlateral displacement of the soft foundation [2]. Althoughgeosynthetics have been widely used in practice to improve
the stability of various geotechnical projects, thereinforcement mechanisms are still far from clear [3], so
designing an embankment on a soft soil still raises several
concerns related to the weak geotechnical properties of thesoft soil. In recent years, there have been a series ofcentrifuge
tests
conducted
onscaled-down
geosynthetics-reinforce structures to investigate thereinforcement mechanisms.
This paper deals with the study of settlement anddeformation behavior of reinforced and unreinforced
embankments subjected to accelerate gravity conditions insmall beam centrifuge at the department of geology of
Universiti Kebangsaan Malaysia (UKM). Series ofcentrifuge tests have been conducted by varying thecharacteristics of foundation soil material.

Reduced scale models are economical and timesaving
alternatives to full scale tasting. Centrifuge model testing is
an example of small-scale models and is a useful tool in theinvestigation of geotechnical problems because of its ability
to reproduce the same stress levels in a small-scale model as
those present in a full-scale prototype [2].
The focus of this paper is on geotextile reinforced
embankment and its deformation behavior is highly
influenced by the presence of prototype stress conditions
and it is possible to achieve those levels of stresses in thelaboratory with the help of centrifuge modeling technique[4].

1

PhD Candidate, department of civil engineering, University Teknologi
Malaysia
2
Associate professor, department of civil engineering, University
Teknologi Malaysia

Centrifuge Modeling

2.1 Description of small geotechnical centrifuge
Centrifuge tests were carry out with small geotechnicalcentrifuge apparatus at department of geology, faculty of
Science and technology FST, national university of
Malaysia (UKM), under the supervision of associatedprofessor Dr. Wan Zuhairi Wan Yaacob.
The apparatus is a small-size beam-type centrifuge,
designed to allow centrifuge testing of soil package up to
6Kg with a maximum rotational speed of 500rpm and canaccelerate up to 140 in units of times gravity at an effective
radius of 0.5 m. The capacity of the small geotechnicalcentrifuge is 0.84 g-ton machine (140 g x 0.6 kg). For
observing the performance of model during the test, thecentrifuge apparatus is equipped with colored digital video
camera and high-speed stroboscope. The recorded ofmoving images can be stored directly into the computer.
Real time video images can also be observed on the LCD
monitor installed in the operation room.
An overall view of centrifuge apparatus and its equipment
is shown in Fig. 1.

637

2.3 Centrifuge models
In this research study, five distinct models base on
different conditions of embankment/foundation anddifferent conditions of reinforcement are tested incentrifuge apparatus.
Case I: Stiff embankment on soft soil foundation(Clay/Kaolin).

Case II: Unreinforced soft soil embankment (Clayeysand) on stiff soil
foundation (Compacte
Case III: Unreinforced soft soil embankment (Clayey sand)on soft soil
foundation (clay/kaolin).
Fig. 1. Overall view of small geotechnical centrifuge
apparatus.

Case IV: Reinforced soft soil (clay/kaolin) embankment on
stiff
soil(Compactedonfoundati sand).

2.2 Test Box
The soil sample box with dimensions of 28cm (length) x
10cm (width) x 19.5cm (height) was placed in a strongbox
that was fixed to the centrifuge arm. The strongbox
constructed with aluminum walls and Perspex at the front toallow the sample to be seen during the test. To adjust thebalance during the flight a counterweight was placed at the
opposite centrifuge arm (Fig. 2).The front wall Perspex
allows viewing and measuring the settlements anddeformations of models after each test. To monitor the soil
settlement during testing, a sketch of embankment andfoundation layers were drawn on the front surface of themodel specimens [5].

Case V: Reinforced soft soil (clayey sand) embankment on
soft soil
foundation (clay/kaolin).
The general condition of each case is listed in Table II. In
four models clayey-sand is used as fill material and in one
model a wooden embankment is used to simulate a stiff
rigid embankment. Compacted sand was modeled as hardsoil foundation while the consolidated kaolin was modeled
as soft soil foundation. Three models are unreinforced andtwo models are reinforced with textile.
The scaled-down centrifugal embankment has a height of
5cm, a crest width of 14cm and slope of 1V to 1H,
underlain by a soft kaolin foundation of 7cm thickness. Due
to inherent symmetry about the centerline, only one half ofthe structure was modeled. Sectional view of model isshown schematically in fig. 3.
Table II. Condition of embankments and foundations incentrifuge tests
NO. Foundation
Fill Material
Reinforcement
1
2
3
4
5

KaolinSand
KaolinSand
KaolinWooden block
Clayey-Sand
Clayey-Sand
Clayey-Sand
Clayey-Sand

TextileTextileFig 2. Small-sized beam- type centrifuge
Details of the small geotechnical centrifuge are shown in
Table 1.
Table I. Specification of small geotechnical centrifuge
Parameter
Value
Package volume
290mm x 100mm x 195mm
Package mass
6 kg
Drive Power
600 W
Rotational speed
80 to 500 rpm
Radius at base package
0.5 m
Acceleration
3 to 140 g
Centrifuge capacity
0.84 g-ton

Fig. 3. Sectional view of embankment model

638

3.

Model Preparation

3.1 Foundation Material
For soft foundation models, kaolin was mixed with anamount of water to obtain the moisture content of 31%(between optimum moisture content and liquid limit) andthen placed in the test box to a thickness of a little bit higher
than 7 cm. The kaolin layer in the test box was rotated using
the centrifuge rotation speed of 50g for about 1 hour toconsolidate and produce homogenous soft soil. At thiscondition, the undrained shear strength of kaolin was foundto be 14 kPa, tested by mini vane shear equipment.
For models with stiff foundation soil, dry sand was poured
into sample box using raining technique and then rotatedabout 30 minutes with the speed of 50g. Table III and IV
show the properties of kaolin and sand. The properties of
the geotextile were obtained from tensile strength test at
department of civil engineering, University Teknologi
Malaysia (UTM) and are given in table V.
Table III. Properties of kaolin
Parameter
Unsaturated unit weight (kN/m3)
Saturated unit weight (kN/m3)
Young’s modulus (kN/m2)
Poisson’s ratio
Cohesion (kN/m2)
Friction angle
Optimum moisture content (%)
Liquid limit, LL (%)
Plastic limit, PL (%)
Plasticity index, I p (%)
Table IV. Properties of sandParameter
Unsaturated unit weight (kN/m3)
Saturated unit weight (kN/m3)
Young’s modulus (kN/m2)
Poisson’s ratio
Cohesion (kN/m2)
Friction angle
Table V. Properties of textile
Parameter
Thickness (mm)Tensile stress (kN/m)Elasticity modulus (Mpa)
Tensile strain (%)

Value
14
17
1000
0.33
5
15
16.5
38
29.7
8.3

Fig. 4. The state of placed kaolin, before performing thecentrifuge test

Fig. 5. The state of placed kaolin, after performing thecentrifuge test

Value
16
20
10000
0.3
1
33

3.2 Fill Materials
In the first model, a wooden block with the dimension ofembankment was used to simulate a rigid, stiff
embankment. For other models oven-dried kaolin mixed
with oven-dried sand to provide clayey sand with 20%
kaolin and 80% sand. Properties of clayey-sand are shownin table VI. To prepare the embankment in the centrifugebox, a specific mold is used on the foundation soil andembankment was constructed by spreading the clayey sandinto the mold and compacted in three layers by wood rod tosimulate a similar compactive effort for all models.

Value
0.17
6.1
65.3
53.4

Fig. 4 and 5 show the kaolin placement before and after
centrifuge test, respectively.

Table VI. Properties of Clayey-Sand
Parameter
Value
Unsaturated unit weight (kN/m3)
16
Saturated unit weight (kN/m3)
19Young’s modulus (kN/m2)
6000
Poisson’s ratio
0.3

639

Fig.6 and Fig.7 show the clayey-sand soil embankments onsand and kaolin foundation before the centrifuge test,
respectively.

Table VII. Gravity acceleration and time maintained ofcentrifuge testTimeω (rpm)Gravity (g)Maintained
G=(1.18×10-3)×r*ω2(minute)
5
120
8.5
5
185
20
5
225
30
15
250
37
10
285
48
20
300
50
5. Results
Case II: Fig. 8 and 9 show the deformation behavior ofmodel II (unreinforced embankment on stiff foundation)
after 60 minutes centrifugal rotation.

Fig. 6. Embankment on compacted sand foundation

Fig 8. Deformation behavior of model II (unreinforced
embankment on stiff foundation)
Fig. 7. Embankment on soft kaolin foundation
4. Centrifuge Test Methodology
After preparing the model in sample box, the model
weighted and then the sample box was fixed to thecentrifuge arm. To balance the system, a counterweightequal to weight of sample added to opposite side of thecentrifuge arm. To reduce the side friction, the sides of thestrongbox were lubricated by applying silicon gel, which
was especially necessary as only half of the embankment
was modeled.
The model was initially run at low acceleration. The models
were subjected to a progressively increasing centrifugal
acceleration limited to 50 g in about 60 minutes. At certain
accelerations the test was maintained constant for certain
times and then again the acceleration continued to increase.
These certain accelerations and time maintained is shown in
Table VII.

Fig. 9. Deformation behavior of model II (unreinforced
embankment on stiff foundation)
The measured settlement on the top of embankment after
the centrifuge test is about 3 mm.
Case III: the side view deformation behavior of model III

640

(unreinforced embankment on soft foundation) aredisplayed in Fig 10, 11 and 12 after 60 minutes centrifugal
rotation.

Fig 10. Deformation behavior of model III (unreinforced
embankment on soft foundation)

about 7 mm.
Case IV: Fig 13 and 14 show the deformation behavior ofmodel IV (reinforced embankment on stiff foundation) after
60 min centrifugal rotation.

Fig. 13. Deformation behavior of model IV (reinforced
embankment on stiff foundation)

Fig. 11. Deformation behavior of model III (unreinforced
embankment on soft foundation)
Fig. 14. Deformation behavior of model IV (reinforced
embankment on stiff foundation)
The measured settlement on the top of embankment after
the centrifuge test is about 0.5 mm.
Case V: the side view deformation behavior of model V(reinforced embankment on soft foundation) are displayedin Fig 15, 16 and 17 after 60 minutes centrifugal rotation.

Fig. 12. Deformation behavior of model III (unreinforced
embankment on soft foundation)
The measured settlement on the top of embankment is641

As expected deformations of model 3 (soft embankment fillmaterials on soft foundation materials) was more than othermodels. To reduce deformations, the embankment was
reinforced with textile. The results show that surface
settlement and deformations of reinforced models are
reduced about 71% for embankment on soft foundation and
83% for embankment on stiff foundation.
As shown in figures the type of deformation for
unreinforced embankment on soft soils (model 3) is deeprotational slide type but for other models is shallowrotational slide type. This change is due to increasing thesoil strength because of reinforcing the soil with textile(model 5) or using stiff foundation at models 2 and 4.
Fig. 15. Deformation behavior of model V (reinforced
embankment on soft foundation)

The effect of side friction between soil and sidewall on
deformation behavior is significant and should beconsidered. Efforts should be made in the centrifuge work
to minimize the effects of side friction by using somemartials like silicon grease to reduce the friction
Acknowledgment
The author would like to thank to associate professor Dr.
Wan Zuhairi Wan Yaacob for his kind guides on centrifuge
test process.

Fig. 16. Deformation behavior of model V (reinforced
embankment on soft foundation)

Fig. 17. Deformation behavior of model V (reinforced
embankment on soft foundation)
The measured settlement on the top of embankment after
the centrifuge test is about 2 mm.
1.

CONCLUSION

Five centrifuge tests were performed based on different
conditions of foundation strength and embankment
reinforcement at the small centrifuge equipment of UKM.
The mini-centrifuge equipment of UKM, does not enable acomprehensive study of reinforced embankments. Thisrefers obviously to the capacity of the centrifuge. The
payload limitation compels one to choose problems withlesser complexities. The size of the strongbox is designed totake into account this limitation and hence make it possible
to study a slope of 1:1 only. However, this apparatus has
enabled the qualitative behavior of the phenomenon to bestudied.
642

References[1] Koerner, R.M., “Designing with geosynthetics” 5thedition. Ed. Pearason Education, 2005.
[2] Sharma, J.S., Bolton, M.D. “Centrifuge modelingof an embankment on soft clay reinforced withgeogrid.” Geotextile and geomembrane. 14 (1):
1-17, 1996.
[3] Chen, J. F., Yu, S.B. and Han, J., “Numericalmodeling of a reinforced embankment based on
centrifuge test dimensions.” ASCE GeoHunan
International Conference, 2009.
[4] Viswanadham, B.V.S., Mahajan, R., “Modelling of
geotextile reinforced highway slopes in a
geotechnical centrifuge” Geotechnical Engineering
for Transportation Projects: Proceedings of
Geo-Trans, ASCE, 637-646, 2004.
[5] Muhardi, “Pulverized fuel ash as structural fill forembankment construction.” 2011.

Centrifuge Model Tests and Finite Element Analyses on Seismic Behavior of
Quay Walls Backfilled with Cement-Treated Granular Soils
Hayano, K.1, Morikawa, Y.2, Fukawa, H.3, Takehana, K.4 and Tanaka, S.1
Yokohama National Univ., 2Port and Airport Research Institute, 3Toa Corporation, 4Geodesign, Japan

1

ABSTRACT
Dynamic centrifuge model tests and finite element analyses (FEA) were conducted to investigate the seismic behaviorof quay walls backfilled with cement-treated granular soils (CTGS). In particular, the effects of the CTGS fill depth andfill range on seismic behavior were investigated. The centrifuge model tests showed that no liquefaction was generated
in the CTGS backfills. The quay wall’s horizontal displacement induced by the seismic loading decreased withincreases in the CTGS fill depth because the earth pressure acting on the quay wall was reduced. In addition, awedge-shaped CTGS fill was found to be effective at reducing the horizontal displacement of the quay wall. However,
the results of the dynamic FEA indicate that the wedge angle of the CTGS fill should be carefully designed.
Keywords: Ground improvement, Cement-treated granular soil, Centrifuge model test, Finite element analyses

Attempts have been made recently to reuse soft soils obtained
from dredging conducted near ports and harbors. The
methods developed so far include dehydration, segregation,and cement-treated techniques. Disadvantages as well as
advantages have been found for each method. For example,
dehydration and segregation methods usually take a long time
to stabilize a large amount of dredged soils. Cement-treatedmethods take a shorter time compared to dehydration andsegregation methods. However, cement-treated materials
show brittle behavior; cracks are generated when thefoundation layers show differential settlement owing toconsolidation.
The focus of this study is on a granulating technique, which
is used for overcoming these difficulties. The method
converts the dredged soils with their high water content togranular materials by adding cement and polymer. To date,
the authors’ research group has conducted fundamental
studies on the geotechnical characteristics of cement-treatedgranular soils (CTGS) [1], [2]. These studies established the
mixture design and the production method to obtain CTGS at
low cost. Based on these findings, Dong et al. (2011)
performed a series of laboratory tests to investigate thephysical and mechanical properties of CTGS[3]. It was
concluded that CTGS is a lightweight material because theparticles include many voids. Therefore, it can be expectedthat CTGS would reduce the earth pressure acting on
retaining walls if used as backfill soils. However, theapplicability is not fully understood at present. In particular, it
is important to clarify the effects of the CTGS fill depth andfill range on the seismic behavior of a quay wall.
Therefore, in this study, a series of dynamic centrifuge
model tests and finite element analyses (FEA) wereconducted to investigate the seismic behavior of quay walls
backfilled with CTGS.

Figure 1 Cement-treated granular soils (CTGS) prepared forcentrifuge model tests
100
80

Percentage finer (%)

1. INTRODUCTION

60
40
20

CTGS
Kawasaki clay
Soma sand

0
1E-3

0.01

0.1

1

Particle diameter (mm)

10

100

Figure 2 Particle size distributions of CTGS, Kawasaki clay,
and Soma sand643

2. CENTRIFUGE MODEL TESTING METHOD
2.1 Materials
CTGS were produced in the following manner for centrifuge
model tests: First, the water content of a dredged soil called
Kawasaki clay was adjusted to 60%. Then, the soil was mixed
with a small amount of polymer with a weight of 0.1% of theclay. Owing to the effect of the polymer, the plasticity of the
mixture became lower. After the plasticity change wasobserved, Ordinary Portland Cement was added and the
mixture was stirred for about 10 min. Granular particles(shown in Fig. 1) gradually appeared because of the effect of
cementation. The weight of the cement added to the mixture
was 5% of the clay. Finally, the granular particles thus
obtained were cured for more than 28 days.
Figure 2 shows the particle size distributions of the CTGS
and Kawasaki clay, as well as that of Soma sand, which was
also prepared for the centrifuge model tests. It is clear that
Kawasaki clay was converted to a gravel-type soil (CTGS) by
the granulation process.
2.2 Model preparation
Four models (denoted by Cases 1 through 4) were prepared tophysically simulate a quay wall, as shown in Fig. 3. The quay
wall consisted of a foundation layer, a caisson, and backfill
soils. As seen in the figures, CTGS and Soma sand were usedas the backfill soils. The foundation layer of each model was
made with Soma sand of 90% relative density and had a
2.5-m thickness at prototype scale.
The sandbox used for Cases 1 and 2 had dimensions of 410
mm (height) by 556 mm (width) by 200 mm (length), whereasthat used for Cases 3 and 4 had dimensions of 512 mm(height) by 710 mm (width) by 200 mm (length) at model
scale.
As seen in Figs. 3(a) and (b), the caisson in Cases 1 and 2
was backfilled with two horizontally layered soils. In Case 1,
the overlying layer was filled with CTGS that was 1.8 m thick(20% of the total thickness of the backfilled ground) at
prototype scale. The underlying layer was Soma sand thatwas 7.2 m thick. Case 2 was similar, but the overlying layer
was 6.75 m thick (75% of the total thickness of the backfilled
ground) and the underlying layer was 2.25m thick.
In Cases 3 and 4, the CTGS was placed in a wedge shapewith one edge of the wedge adjacent to the caisson. In Case 3,
the right angle of the wedge was at the upper left corner of thebackfill, as shown in Fig. 3(c). The remaining part outside thewedge was filled with Soma sand. In contrast, the right angleof the wedge in Case 4 was in the lower left corner of thebackfill, as shown in Fig. 3(d). The remainder was filled with
Soma sand.
In all cases, first, Soma sand was poured using the air
pluviation method, resulting in a foundation layer of relative
density of 90%. Next, in Cases 1, 2 and 3, Soma sand wasfilled using the air pluviation method to become about 50 %relative density. Then, a small shovel was used to place
CTGS in the sandboxes at zero height, such that a loose state
was achieved. In Case 4, CTGS was filled, followed by thefill of Soma sand of 50% relative density.
In each case, two displacement transducers (denoted by D1
and D2 in the figures) were set at the seaward side of thecaisson to measure its horizontal displacement. The other two
displacement transducers (denoted by DT1 and DT2) were set
above the backfill soils to measure the ground surface
settlement. Pore water pressure gauges and accelerometers
were installed inside the backfill soils at prefixed positions, as
shown in the figures.

(a) Case 1(b) Case 2(c) Case 3(d) Case 4
Figure 3 Schematic images of models prepared for centrifuge
model tests644

To evaluate the earth pressure distribution acting on the
caisson, four load cells (denoted by LC1 through LC4) werebuilt in the caisson. A rigid rectangular plate that directly
contacted the backfill soils was connected to each load cell.
The plate surface was set to be smooth in this study. A small
space was allowed between the plates, so that they would notinterfere with each other.
In preparing the models, efforts were made to minimize thefriction between the inside of the box and the model soils byinserting thin, smooth films.

(At prototype scale)

160

Case1
Case2
Case3
Case4

140
120
100

3.1 Horizontal displacement of the caisson and theground surface settlement
Figure 4 shows the relationship between the maximum
amplitude of the input acceleration and the cumulative
horizontal displacement of the caisson for each case. The
horizontal displacement was measured at the upper part of theseaward side of the caisson (see Fig. 3). The relationships
obtained from the previous study [6], [7] were also shown in Fig.
4.
The previous study included three centrifuge model tests,
denoted here by loose sand (LS), dense sand (DS) and CTGS
(CT). The LS model was backfilled solely by Soma sand of
50% relative density, while the DS model was backfilled only
by Soma sand of 95% relative density. The CT model was
backfilled solely by loose CTGS. The caisson used in the LS,
DS, and CT tests was the same as that used in this study.
Figure 4 shows that LS had the largest horizontal
displacement at the higher accelerations. This was because
the Soma sand was liquefied, and the earth pressure acting on
the caisson increased as compared to other models. This will
be discussed in detail in sections 3.3 and 3.4.
The caisson’s horizontal displacement in Case 1 was close
to that of LS, despite the fact that the upper part (20% of thetotal thickness) of the backfill ground consisted of CTGS. On
the other hand, the caisson did not show a large horizontal
displacement in Case 2. The horizontal displacement wasclose to that of DS. The caisson showed the smallest
displacement in the CT case. These facts indicate that thecaisson’s horizontal displacement was reduced as the fill
depth of the CTGS increased.
Looking at the results of Cases 3 and 4 in Fig. 4, thecaissons show a horizontal displacement close to that of Case
2. This fact suggests that CTGS filled in the shape of a wedge
or a reversed wedge can be effective at reducing the caisson’s
horizontal movement. The total CTGS volume required forCase 3 or 4 was lower than that required for Case 2.

Results from the previous study
LS(Loose Soma sand)DS(Dense Soma sand)
CT(Loose CTGS)

80
60

Ground surface settlement of backfill soil
at a location 4 m* from the caisson (cm)Horizontal displacement of the caisson
measured at the upper part (cm)

2.3 Centrifugal acceleration and dynamic shaking
A geotechnical centrifuge facility known as Mark II at thePort and Airport Research Institute was used in this study.
The beam radius of the centrifuge is about 3.8 m and the
maximum loading mass is 2.76 ton. Detailed information canbe found in Kitazume and Miyajima (1995) [4].
Centrifugal acceleration and dynamic shaking wereconducted in the following manner: First, after each model
was prepared, it was set on the shaking table installed on thecentrifuge platform. The model was then accelerated up to 20
G by the centrifuge. At the 20-G acceleration field, viscous
fluid was infiltrated into the ground from the bottom tosaturate the soils. The high saturation degree of the soils was
obtained as proposed by Okamura and Kitayama (2008) [5].
The fluid’s viscosity was set to be 50 times that of water tosatisfy the similitude of the pore water dissipation of theground. Then, the centrifugal acceleration was increased up
to 50 G, and dynamic shaking was applied six or seven times.
In each shaking period, 20 cycles of the sinusoidal waves of
2-Hz frequency at prototype scale were applied. The
maximum amplitude of acceleration applied to the ground
was increased stepwise up to about 400 Gals.

3. CENTRIFUGE TEST RESULTS

40
20
0

0

100

200

300

400

Maximum amplitude of input acceleration (Gal)

Figure 4 Relationship of input acceleration and caisson
horizontal displacement(At prototype scale)

140

Case1
Case2
Case3
Case4

120
100

Results from the previous study
LS(Loose Soma sand)DS(Dense Soma sand)
CT(Loose CTGS)

80
60
40
20
0

*4.15m for Cases 3 and 4

0

100

200

300

400

Maximum amplitude of input acceleration (Gal)

Figure 5 Relationship of input acceleration and backfill
surface settlement645

Figure 5 shows the relationships between the maximum
amplitude of the input acceleration and the cumulative
settlement of the ground surface. The surface settlement wasmeasured at locations 4.0 or 4.2 m from the caisson edge (seeFig. 3). As shown in the figure, LS has the largest groundsurface settlement, owing to the liquefaction of the Soma
sand.
Wedge-shaped CTGS fills were a common characteristic
between Cases 3 and 4; however, Case 3 shows a notably
smaller ground surface settlement than Case 4. The groundsurface settlement of Case 3 was close to that of DS. The
reason why a small ground surface settlement was observedin Case 3 will be discussed in section 4.3.
G.L.-1.5m
G.L.-4.9m
G.L.-7.2m
Input acceleration

Acceleration (Gal)

400
300
200
100
0
-100
-200
-300 (At prototype scale)
-400
6.0
6.5

7.0
Elapsed time (s)

7.5

8.0

Acceleration (Gal)(a) Case 1
400
300
200
100
0
-100
-200
-300 (At prototype scale)
-400
6.0
6.5

G.L.-3.0m
G.L.-5.3m
G.L.-8.5m
Input acceleration

7.0
Elapsed time (s)

7.5

8.0

Acceleration (Gal)(b) Case 2
400
300
200
100
0
-100
-200
-300 (At prototype scale)
-400
6.0
6.5

G.L.-3.3m
G.L.-6.2m
G.L.-8.2m
Input acceleration

7.0
Elapsed time (s)

7.5

8.0

(c) Case 3
Figure 6 Time histories of ground response and input
accelerations
3.2 Ground response acceleration
Figure 6 shows the time histories of the accelerations
recorded in the backfill soil in Cases 1, 2, and 3. The data
presented here were obtained in each case during the 6–8 sthat elapsed after the start of the third shaking period. The
ground acceleration was measured at locations 4.0 or 4.2 m
from the caisson. Acceleration time histories were notobtained in Case 4 owing to connection errors with theaccelerometers.
The ground accelerations were attenuated in each model as
seen in the figures. In particular, the maximum amplitude ofaccelerations obtained at ground depths of 1.5 and 4.9 m inCase 1 were smaller than those obtained in Cases 2 and 3. It isbelieved that the stiffness of Soma sand was considerably
reduced by liquefaction and that the shear stress transmitted
in the ground was attenuated in Case 1.
Phase delays were also observed in the ground acceleration
waves. The wave phase differences between the acceleration
obtained at a depth of 1.5 or 4.9 m and the input acceleration
were about 0.2–0.3 s in both Cases 2 and 3.
3.3 Excessive pore water pressure
The ratio of the maximum excessive pore water pressure to
the effective overburden pressure ∆u/σ v ' was evaluated basedon measurements made with the pore water pressure gauges.
Figure 7 shows the relationships between ∆u/σ v ' and theground depths for Cases 1–4. The excessive pore waterpressure was measured at locations 12.2 or 13.5 m from thecaisson. The data presented here were obtained at the third,
fourth, and fifth shaking periods (periods 3–5) in each case.
The calculated effective overburden pressure was also shownin the figures.
As shown in Fig. 7(a), the values of ∆u/σ v ' obtained at
ground depths of 4.4 and 7.0 m were close to 1.0, whichindicated the liquefaction of loose Soma sand in Case 1. In
addition, the values of ∆u/σ v ' obtained at the Soma sandlayers in Cases 3 and 4 were close to 1.0, as seen in Figs. 7(c)and (d). This was also because of liquefaction.
On the other hand, ∆u/σ v ' obtained at a ground depth of 1.4
m in Case 1 was about 0.3. Small ∆u/σ v ' values were also
observed at ground depths of 3.1 and 5.0 m in Case 2, 1.3 m inCase 3, and 8.3 m in Case 4. These values were all obtainedwith the pore water pressure gauges installed in the CTGS.
These findings verified that liquefaction was not observed in
CTGS, even though the CTGS fill was loose and at shallow
ground depths. The low liquefaction potential of CTGS canbe explained by the high permeability characteristic of CTGS,
which is suggested by the grain-size distribution shown in Fig.
2.
3.4 Dynamic earth pressure distribution
Figure 8 shows the dynamic earth pressure distribution acting
on the caisson in each case. The dynamic earth pressure
presented here was the minimum–maximum range obtained
during period 3. The dynamic earth pressure was calculated
by subtracting hydrostatic pressures from the total earthpressures measured with the load cells.
The figures show that the dynamic earth pressure at ground
depths of 3.2, 5.3, and 7.4 m in Case 1 was higher than thoseobtained in the other cases. Similarly, a higher dynamic earthpressure was observed at a ground depth of 7.4 m in Case 2.
These higher pressures were induced by the effect of theliquefaction of the Soma sand.
On the other hand, the dynamic earth pressure obtained at a
ground depth of 1.1 m in Case 1 and at ground depths of 1.1,
3.2, and 5.3 m in Case 2 were not so high. In addition to thefact that no liquefaction was generated in CTGS, it is believedthat the CTGS layers in Cases 1 and 2 induced a small

646

Excessive pore water pressure ratio, ∆u/σ'
0.5

1.0

0

Measured at locations 13.5 m from the caisson

CTGS

2

Ground depth (m)Ground depth (m)0.00Soma sand4
6 Effective overburden pressure
Excessive pore water pressure ratio
Period 3(Max. accel.:207Gal)
Period 4(Max. accel.:310Gal)
Period 5(Max. accel.:369Gal)

8
0

20

40

0.5

Ground depth (m)Ground depth (m)Measured at locations 13.5 m from the caisson

Excessive pore water pressure ratio
Period 3(Max. accel.:212Gal)
Period 4(Max. accel.:312Gal)
Period 5(Max. accel.:371Gal)

CTGS
Soma sand

8
0

20

40

60

Effective overburden pressure, σv'（kN/m2）

0

10

20

30

40

Dynamic earth pressure (kN/m2)

50

60

0

Measured at locations 13.2 m from the caisson

4
6 Effective overburden pressure

40

2
4
6
CTGS

8

Excessive pore water pressure ratio
Period 3(Max. accel.:188Gal)
Period 4(Max. accel.:354Gal)
Period 5(Max. accel.:349Gal)

20

Soma sand

1.0

2

0

CTGS

(b) Case 2

Ground depth (m)Ground depth (m)

6

-10

80

CTGS
Soma sand

8

-10
60

Effective overburden pressure, σv'（kN/m2）

80

0

10
20
30
40
50
Dynamic earth pressure (kN/m2)

60(c) Case 3(c) Case 30
Excessive pore water pressure ratio, ∆u/σ'
0.5
1.0

Ground depth (m)0.00Ground depth (m)4Excessive pore water pressure ratio, ∆u/σ'
0.5

60

2

8

(b) Case 2
0.0
0

10
20
30
40
50
Dynamic earth pressure (kN/m2)

0

1.0

6

0

(a) Case 1

Excessive pore water pressure ratio, ∆u/σ'

4

6

-10

Effective overburden pressure

2

4

80

(a) Case 1
0.0

Soma sand

860

Effective overburden pressure, σv'（kN/m2）

0

CTGS

2

Measured at locations 13.2 m from the caisson
Excessive pore water pressure ratio
Period 3 (Max. accel.:205Gal)
Period 4 (Max. accel.:295Gal)
Period 5 (Max. accel.:332Gal)

2
4

Effective overburden pressure

6

0

4
6

-10
20

CTGS

8

Soma sand
CTGS

8

2

40

60

Effective overburden pressure, σv'（kN/m2）

0

10
20
30
40
50
Dynamic earth pressure (kN/m2)

60(d) Case 480(d) Case 4
Figure 7 Depth distributions of excessive pore water pressure
ratio and effective overburden pressure

Figure 8 Depth distributions of dynamic earth pressure
distributions at caisson wall

647

Table 1 Strength parameters used in dynamic FEA based on
CD triaxial tests

Soma sand(D r = 50%)
Soma sand(D r = 90%)
CTGS
120

CTGS
Water

CaissonSoma sand（Dr=50%）

Soma sand（Dr=90%）

Figure 9 Finite element meshes adopted to simulate Case 1 indynamic FEA(kN/m2)

8.5

0.6

36.1

9.5

8.6

40.4

4.3

0.0

26.6

φd

50

σc' =50kPa

40

Fitting curves for finite numerical analyses

80

30

60

20

40

10

20
0
1E-3

0.01

0.1

Single amplitude of shear strain, γSA (%)

1

0

Figure 10 Comparison of laboratory and FEA results forequivalent shear modulus and hysteric damping ratio as afunction of shear strain
0.20
Shear stress ratio, σd/(2σ0')

4.1 Software and numerical analysis conditions
The centrifuge model test results indicated that thewedge-shaped fill of CTGS in Case 3 was effective at
reducing the caisson’s horizontal displacement and thebackfill ground surface settlement when the quay wall was
subjected to the dynamic loading. However, owing to thelimited conditions of the centrifuge model tests, theappropriate range of the wedge-shaped zone was not wellunderstood. Therefore, dynamic FEA were conducted toobtain the effect of the wedge-shaped zone range (inparticular, the effect of the wedge angle) on the caisson’s
horizontal displacement.
Dynamic FEA software named FLIP was used in this study.
The software is widely used for seismic designs of port andharbor structures in Japan (e.g., Iai and Kameoka, 1993[8], Iaiet al., 1998[9]). First, the analyses using FLIP were focused on
simulating the effect of the fill depth of CTGS on the
caisson’s horizontal displacement. The centrifuge models(Cases 1 and 2 in this study and LS and CT in the previous
study) were simulated. An analysis was also conducted tosimulate the model in which the upper half of the backfill was
CTGS.
As an example, the finite element meshes used to simulate
Case 1 are shown in Fig. 9. The scales of the foundation layer,
the caisson, and the backfill layers in the figure were set to be
the same as those of Case 1 at the prototype scale (see Fig. 3).
For the dynamic acceleration, sinusoidal waves that were thesame as those of the centrifuge model tests were introduced
into the FEA. The maximum amplitude of the acceleration
was adjusted to 200 Gal.

cd
(kN/m2)

CTGS from triaxial test
Soma sand (Dr=90%) from torsional test

100

4 FINITE ELEMENT ANALYSES

Submerged unit
weight, γ’ (kN/m3)

Hysteric damping ratio, h (%)

Material

Equivalent shear modulus, Geq (kN/m2)

dynamic earth pressure because of their light weight. This
light weight characteristic is attributable to the fact that manysmall voids were included in the particles. Please see Dong etal. (2011) [3] for further information related to this issue.
The dynamic earth pressures obtained at ground depths
of 3.2 and 5.3 m in Cases 3 and 4 were slightly higher thanthose in Case 2. This was probably induced by theliquefaction of the Soma sands filled behind the CTGS. Thissuggested that the effect of liquefaction of Soma sand filled
behind the CTGS might be higher if the fill zone range of awedge of CTGS became smaller. This matter will be
investigated in detail in the following sections.

Soma sand (Dr=50%)

0.15
0.10

Undrained cyclic triaxial tests
Relationship estimated for finite element analyses

0.05
0.00

1

10
Number of cyclic loading, Nc

100

Figure 11 Comparison of liquefaction resistance curves
obtained from laboratory tests with those estimated based on
FEA
The material parameters related to the strength propertiesof CTGS and Soma sands were determined based on the
consolidated drained triaxial tests are listed in Table 1, andthose related to the deformation properties were determinedbased on cyclic triaxial tests or cyclic torsional tests, as shownin Fig 10. Laboratory experiments were conducted according
to JGS 0524-2009, JGS 0542-2009 and JGS 0543-2009.
The liquefaction parameters necessary for simulating theliquefaction behavior of Soma sand were evaluated based on
the cyclic undrained triaxial tests (JGS 0541-2009), as shownin Fig. 11. The CTGS were treated as nonliquefiable soils

648

50
Centrifuge model test
Dynamic finite element analyses

40
30

parameters and the constitutive models, analyses wereconducted to investigate the effect of range of thewedge-shaped CTGS fill zone (in particular, the effect of thewedge angle) on the caisson’s horizontal displacement. The
wedge angle θ was defined as the inclination angle of thewedge plane from the horizontal plane. Typically, as shownin Fig. 13, the wedge angle θ was set to 0°, 30°, 45°, 60°, and
90° in each analysis. The scales except for the wedge angle
were designed to be as same as those of the centrifuge model
tests.
Horizontal displacement of the caissonat the upper part of the sea side (cm)Horizontal displacement of the caissonat the upper part of the sea side (cm)based on the centrifuge model test results. Detailed
information on the input parameters required for FLIP and theconstitutive material models can be found in Morita et al.
(1997) [10].

2010(At prototype scale)
0
0.0
0.2
0.4
0.6
0.8
Normarized fill depth of CTGS

1.0

Figure 12 Comparison of centrifuge model tests and FEA
results for caisson horizontal displacement as a function of fill
depth

70
60

(At prototype scale)

50
40
30
20

Caisson(Case3)

10
0

CTGS
Water

(LS)

(CT)

0

Finite element analyses
Centrifuge model tests
30
60
Wedge angle, θ (°)

90

Figure 14 Comparison of centrifuge model tests and FEA
results for wedge-angled CTGS fill and caisson horizontal
displacement

Soma sand（Dr=50%）

θSoma sand（Dr=90%）

Figure 13 Example of finite element meshes adopted torepresent wedge-shaped CTGS fill (θ = 60°)
4.2 Numerical analysis results on the effect of the filling
depth of CTGS
Figure 12 shows the relationship between the fill depth of
CTGS and the horizontal displacement of the caisson. The fill
depths of CTGS were normalized by the total thickness of thebackfill soil. The horizontal displacements of the caisson
were evaluated at the upper part of the seaward side of thecaisson, the same as for the centrifuge model tests (see Fig. 3).
For comparison, the centrifuge model test results are also
shown in the figure. The figure shows that the FEA results inhorizontal displacements that are close to the centrifuge
model test results. These results suggest that the horizontal
displacement of the caisson was reduced proportionally to thenormalized fill depth.
4.3 Numerical analysis results on the effect of the rangeof the wedge-shaped fill zone
As described in the previous section, the horizontal
displacement of the caisson could be reasonably predicted by
the finite element analyses by introducing the appropriate
parameters into the numerical models. Then, using the same(a) θ = 30°(b) θ = 60°
Figure 15 Maximum shear strain distribution obtained fromdynamic FEA

649

Figure 14 shows the relationship between the wedge angle

θ and the horizontal displacement of the caisson. The

centrifuge model test results were also shown in the figure.
Here, the wedge with the 0° angle represented the backfill
ground solely filled by CTGS (CT) and that of the 90° angle
represented the backfill ground solely filled by Soma sand(LS). It can be seen in the figure that the wedge with a 60°
angle showed the largest horizontal displacement.
Figure 15 shows the maximum shear strain distributions of
the grounds for the wedges with 30° and 60° angles. The
results indicate that higher shear strain was concentrated near
the boundary between CTGS and Soma sand in the 60° wedge
case compared to that in the 30° wedge case. This may
suggest that the range of the wedge-shaped CTGS fill would
be better designed to cover the active failure zone of backfill
ground.
In addition, Figure 15 shows that Soma sand moved in thehorizontal direction toward the caisson after it was liquefied.
This sand appeared to push the CTGS up toward the groundsurface during the movement because the density of theliquefied soil was much higher than that of CTGS. Thismechanism can explain the reason why a small amount ofground surface settlement was observed in Case 3 but not inCase 4 in the centrifuge model tests.
5 CONCLUSIONS
Dynamic centrifuge model tests and finite element analyses(FEA) were conducted to investigate the seismic behavior ofquay walls backfilled with cement-treated granular soils(CTGS). Effects of the fill depth and fill range of CTGS on
the seismic behavior were investigated. The following results
were obtained:
1) The centrifuge model tests showed that no liquefaction was
generated in the CTGS backfills. The low liquefaction
potential of CTGS was due to the highly permeablecharacteristic of CTGS resulting from the coarse grain-size
distribution.
2) The quay wall horizontal displacement induced by theseismic loading was decreased as the CTGS fill depth
increased because the earth pressure acting on the quay wall
was reduced. The small earth pressure was attributed to thelight weight characteristic of the CTGS.
3) The centrifuge model tests also showed that thewedge-shaped CTGS fill was effective at reducing the quay
wall’s horizontal displacement and the ground surface
settlement.
4) The dynamic FEA using FLIP could reasonably give thecaisson’s horizontal displacements, which were close to thoseof the centrifuge model tests. The FEA also indicated that thewedge angle of the CTGS fill zone should be carefully
designed to cover the active failure zone of the backfill
ground.
5) The dynamic FEA for the wedge-shaped fill also showedthat liquefied Soma sand filled behind the CTGS and pushed
the CTGS up toward the ground surface. This contributes tothe fact that a small ground surface settlement was observed
for the wedge-shaped CTGS fill in the centrifuge model tests.

REFERENCES[1] Dong, P. H., Hayano, K., Okusa, Y., Takahashi, H. andMorikawa, Y., “Mechanical Properties of Lean-Mixed
Granular Cement Treated Soil from Consolidated
Drained Triaxial Tests”, Proc. of International
Symposium on Geotechnical Engineering, Ground
Improvement, and Geosynthetics for Sustainable
Mitigation and Adaptation to Climate Change Including
Global Warming, pp. 319–330, 2009.
[2] Dong, P. H., Hayano, K., Takahashi, H. and Morikawa,Y., “Effects of Mixture Design on the Mechanical
Properties of Cement Treat Granulate Soil,” Proc. of
Sixth International Congress on Environmental
Geotechnics, Vol. 1, pp. 594–599, 2010.
[3] Dong, P. H., Hayano, K., Kikuchi, Y., Takahashi, H., andMorikawa, Y., “Deformation and crushing of particles of
cement treat granulate soil”, Soils and Foundations, Vol.
51, No. 4, pp. 611–624, 2011.
[4] Kitazume, M. and Miyazima, S., “Development of PHRIMark II geotechnical centrifuge”, Technical Note of thePort and Harbor Research Institute, No. 817, pp. 1–33,
1995.
[5] Okamura, M. and Kitayama, H., “Preparation of Fully
Saturated Model Ground in Centrifuge and HighAccuracy Measurement of Degree of Saturation, Journalof JSCE, Division C: Geotechnics，Vol. 64，No. 3，pp.
662–671, 2008 (in Japanese).
[6] Morikawa, Y., Takahashi, H., Hayano, K. and Okusa, Y.,“Centrifuge Model Tests on Dynamic Behavior of Quay
Wall Backfilled with Granular Treated Soil,” Proc. of the
8th International Conference on Urban Earthquake
Engineering, pp. 279–284, 2011.
[7] Takahashi, H., Morikawa, Y., Hayano, K. and Okusa, Y.,“Seismic behavior of gravity-type quay wall backfilledwith granular soil”, Japanese Geotechnical Journal, Vol.
6, No. 2, pp. 169–179, 2011 (in Japanese).
[8] Iai, S. and Kameoka. T., “Finite element analysis ofearthquake induced damage to anchored sheet pile quay
walls,” Soils and Foundations, Vol. 33, No. 1, pp. 71–91,
1993.
[9] Iai, S., Ichii, K., Li, H., and Morita, T., “Effective stressanalyses of port structures,” Soils and Foundations,
Special Issue on Geotechnical Aspects of the January 17
1995 Hyogoken-Nambu Earthquake No.2, pp. 97–114,
1998.
[10] Morita, T., Iai, S., Liu, H., Ichii, K. and Sato, Y.,“Simplified Method to Determine Parameter of FLIP”,
Technical Note of the Port and Harbor Research Institute,
Ministry of Transport, Japan, No. 869, pp. 1–36, 1997 (inJapanese).

650Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Response of Piered Retaining Walls to Lateral Soil Movement Based on
Numerical Modeling using Matlab
Siamak Pahlevanzadeh1
1

Griffith School of Engineering, Griffith University, Gold Coast Campus, QLD 4222, Australia; PH (0061)
7-5552 7608; FAX (0061) 7-5552 8803; email: s.pahlevanzadeh@griffith.edu.au
ABSTRACTThe response of individual retaining walls employed as a preferred slope stabilization technique, to lateral loading has
been widely explored in the literature. The present study focuses on the response of piered retaining walls, i.e. retaining
walls supported by a pile foundation system, to lateral soil movement. Numerical analysis results by MATLAB haverevealed wall-soil-pile interaction mechanism with different effective height of wall, spacing, diameter and number ofpiles, using different soil properties, which quantifies the behavior of piered retaining walls owing to lateral soil
movement. Based on a numerical configuration sensitivity study, the relationship between lateral loading and pile/walldeflection, shear and bending moment under different influencing parameters has been established. Optimum effective
parameters are accordingly defined and relevant guidance on choosing appropriate piling configuration is provided,
which can be of practical use in engineering design.
Keywords: Piered Retaining Wall, Lateral Soil Movement, Numerical Modeling, Matlab

1. INTRODUCTION
The problem associated with lateral earth pressure andretaining wall stability is one of the most common in the civilengineering field and a segment of soil mechanics that has
been receiving widespread attention from engineers for a longtime.
The typical structures whose primary or secondary purpose isto resist earth pressures may include various types ofretaining walls, sheet piling, braced sheeting of pits andtrenches, bulkheads or abutments, and basement or pit walls.
These may be self-supporting (e.g., gravity or cantileverconcrete walls) or they may be laterally supported by meansof bracing or anchored ties. The lateral earth pressure depends
on several factors [1]:• The physical properties of the soil
• The time-dependent nature of soil strength• The interaction between soil and retaining structure at
the interface• The general characteristics of the deformation in thesoil-structure composite• The imposed loading (e.g., height of backfill,surcharge loads).
Due to virtually no analytical, theoretical or field testing dataon piered retaining walls, this review will focus on theliterature on wall-soil interaction and the behavior of piles
subjected to a lateral loading. Although the integrated
structure will behave differently, numerical modeling can still
be employed on the basis of how each section would behave

in a traditional situation, i.e. the retaining wall behaves as apiling cap with an additional moment force due to earthpressure.
Although much research has been performed and appreciableadvancement made during the past two centuries regarding
the distribution of earth pressures and on the analysis of awide range of earth-retaining structures, some of the theories
formulated by Coulomb (1776), Rankine (1857) andMononobe-Okabe (1929) still remain as the fundamental
approaches to the analysis of most earth-supporting structures,
particularly for sandy soils. Furthermore, although someresearch data and experience indicate that assumptions
related to pressure distributions on retaining walls, or on the
failure surface of the backfills, are not quite those depicted bythese early investigators, substantial evidence exists that theanalysis and design efforts based on their theories give
acceptable results for most cases of cohesion-less backfills.
The results are significantly less dependable for the morecohesive soils.
Recent analyses have tended to concentrate on numericalmethods, in particular, the three-dimensional finite element
methods. The importance of incorporating interface elements
to simulate possible slippage and separation between the walland soil, and capturing the soil nonlinearity using advanced
constitutive models, has been widely recognized.
Advanced numerical analysis may be appropriate andnecessary for detailed design; generally, the wall-soil
interaction may be addressed with numerical simulations
based on either the finite-element method (FEM), or thefinite-difference method (FDM).
Goh (1993) performed finite element analyses to investigate
the effects of subsoil stiffness, wall stiffness, and wall651

roughness on the lateral earth pressure for concrete cantileverretaining walls, and proposed a modified earth pressure
distribution.
In spite of recent published methods, the tendency today inpractice is to use the values given by Caquot and Kerisel(1948) [2] and Kerisel and Absi (1990) [3].
Benmebarek et al. (2007) [4] concerned with the numericalevaluation of passive earth pressure coefficients for a rigid
rough vertical retaining wall with horizontal ground surface
and limited breadth subjected to translation. They conducted
a numerical study of 3D passive earth pressures induced by
the translation of a rigid rough retaining wall for associativesoils. Using the explicit finite difference code FLAC3D, theincrease of the passive earth pressures due to the decrease of
the wall breadth was investigated. FLAC uses an explicit
finite difference program to study numerically the mechanicalbehavior of a continuous 3D medium as it reaches
equilibrium or steady plastic flow. The explicit Lagrangiancalculation scheme and the mixed-discretization zoning
technique as presented by Cundall (1987) [5], is used in
FLAC to ensure that plastic failure and flow are modeled
accurately.
Numerical methods including both finite element and finite
difference methods have been widely employed to study thewall-soil interaction under active and passive lateral loading
due to soil movements. These methods mainly satisfy thecompatible relationship between the lateral pressure and the
movement. However, relatively little effort is made to assess
the response of the piered retaining walls under the action of
the lateral soil movements.
2. NUMERICAL MODELING
2.1 Method
The numerical model was created by combining two methods(Fig. 1) using C++ script in Matlab:• Finite Difference Method, FDM, (Matlock andReese, 1960) [6] revised by Guo (2003) [7],• Finite Element Method, FEM, reviewed andverified by Chen (1998) [8] as the method ofsuperposition, namely flag pole analysis
Basic mechanics of materials formulae is used to determine
elastic deflection in the retaining wall, boundary conditions tohave a statically determinant cantilever wall which can beused to determine the degree of deflection the wall itself
undergoes.
Numerical modeling using Matlab is designed to calculate
deflection when subject to a variety of parameters, maximum
shear in each pile correlating to the applied load and bending
moment at different depths, to find where the maximum
moment occurs. Numerical modeling is also designed tohighlight the impact of input parameters on the performanceof a piered retaining wall to facilitate the development ofdesign guidelines.

Figure 1 - The numerical model was created by
combining two methods, FDM and FEM
By implementing the popular p-y method, Rankine’s
Coulomb’s and Mononobe-Okabe’s general formulas for acase of static equilibrium using Matlab, results can beobtained to verify the accuracy of existing methods applied toan unfamiliar situation. This method can only be reliable for apile group with S/D of 6 or greater where S, is the spacing
between piles and D, is the diameter of each pile.
2.2 Analysis
The numerical model uses the FDM to determine themaximum bending moment and the angle of rotation in thepiles, and the FEM analysis to determine pile deflection.
Mechanics of materials formulae is also used to determine
elastic deflection in the retaining wall. The method of criticalpile depth was utilized to establish pile flexibility andboundary conditions required for a statically determinantcantilever wall to be used to determine the deflection of thewall.
The type of piered retaining wall modeled, can be either acantilever or gravity wall retaining wall interface, with one or
two rows of piles beneath it. Finite element analysis is used todetermine this behaviour. The retaining wall’s deflection isdetermined with an extension of the P-y method using thegeneral limit static equilibrium case. The force profile andbehaviour of piles are modelled with p-y curves to determine
the elastic-plastic sliding depth and bending moment for a
non-linear loading distribution profile.
The model is developed for a piered retaining wall subjectedto active earth pressure, which behaves as a pile with fixed
head. Piles are subject to passive earth pressure loading case,where soil movement occurs exclusively at and above thebottom of the retaining wall. The Matlab script is split up into
three parts, input parameters, deflection calculation and pile
analysis. It basically calculates the percentage of the loadingtransferred from the wall to the piles due to the lateral soil
movement and then quantifies the behavior of piles due to theload transferred.
2.3 Model Modifications
A variety of modifications have been made to the numerical652

model to account for the wall – pile interaction. It was foundthat the numerical model best suited for a piered retaining
wall which is developed for substantially long piles. In order
to counteract this problem, modifications were made byimposing a relationship between the critical pile depth and thetotal pile length, such that the deflection was notunderestimated.
Basically if a pile decreases in length the deflection will
decrease, past a certain point namely critical pile depth and if
a pile decreases in length below the critical pile depth,
deflection will increase. After satisfactory equations were
derived and verified, numerical results have been produced.
Comparisons and conclusions have been obtained to deduce
the suitability and feasibility of implementing piered
retaining walls. The purpose of these comparisons is todetermine the sensitivity of the numerical model to the
influencing parameters which establishes the performance ofthe piered retaining wall.
2.4 Results and Discussion
Fig. 2 shows the comparison between three different piling
configurations for a concrete retaining wall. It includes:• two piles, diameter of each pile 177mm, S/D>6• Six piles, diameter of each pile 100mm, S/D>6• Ten piles, diameter of each pile 78mm, S/D<6
The maximum shear force per pile due to a certain lateral loaddecreases considerably by increasing the number of piles
from 2 to 6; however increasing the number of piles from 6 to
10 slightly decreases the maximum shear force per pile. The
bending moment per pile at a certain depth also decreases
considerably by increasing the number of piles from 2 to 6;
however increasing the number of piles from 6 to 10 slightly
decreases the bending moment force per pile.

Figure 3 - Numerical comparison between three different
cases of soil properties for a large reinforced concrete
piered retaining wallFig. 3 shows the comparison between three different cases ofsoil properties for a large reinforced concrete piered retaining
wall. These cases are including sand with:• φ =30 ̊ friction angle and 79% relative density• φ =32.5 ̊ friction angle and 84% relative density• φ =35 ̊ friction angle and 89% relative density
The maximum shear force per pile due to a certain lateral loadincreases by decreasing the friction angle and relative densityof soil. The bending moment per pile at a certain depth also
increases by decreasing the angle of friction and relative
density of soil.

Figure 2 - Numerical comparison between three different
piling configurations for a concrete retaining wall

Figure 4 - Numerical comparison between effective wallheights 150mm, 200mm, and 250mm
653

The deflection at top of wall due to a constant lateral loading
increases accordingly by decreasing the angle of friction andrelative density of soil.
Fig. 4 shows the comparison between three different cases ofeffective wall heights for a large reinforced concrete piered
retaining wall. These cases are including piered retaining wallwith:
• h= 150mm, 190mm submerged into the soil
• h= 200mm, 140mm submerged into the soil
• h= 250mm, 90mm submerged into the soil
Where h is the effective wall height. The maximum shear
force per pile due to a certain lateral load increases by
increasing the effective wall height. The bending moment per
pile at a certain depth also increases by increasing theeffective wall height. The deflection at top of wall due to a
constant lateral loading increases accordingly by increasing
the effective wall height.
3 CONCLUSION
To gain a relatively better understanding of the effect ofpiling beneath a simple typical retaining wall, a numericalmodeling and analysis has been conducted using Matlab. The
model utilizes a combination of both FDM and FEM methods
in order to calculate the earth pressure on the wall due toapplied lateral loading, determine the percentage of loadtransferred from wall to piles and quantify the pile behavior
accordingly.
Discussing the improvement of the engineering behavior of asimple typical retaining wall by piling beneath the retaining
wall also reveals that:• There is optimum number of piles to use beneath any
particular retaining wall, since the numerical modelingindicates that the maximum shear force per pile due toa certain lateral load and the maximum bending
moment at a certain depth decrease by increasing thenumber of piles to a certain number and additional
piling will not considerably improve the lateralcapacity. This optimum number of pile can be
determined running the developed numerical model inthe present study.
• Decreasing the friction angle and relative density ofsoil, increases the maximum shear force per pile due toa certain lateral load and also the bending moment per
pile at a certain depth. This conclusion illustrates theeffect of soil properties on the model. The deflection at
top of wall due to a constant lateral loading increases
accordingly by decreasing the angle of friction andrelative density of soil. The numerical model
developed in the present study is able to consider
different soil properties in order to produce
appropriate outcome.
• There is optimum effective wall height, sinceincreasing the effective wall height will result toincrease in the maximum shear force per pile due to acertain lateral load and also the bending moment per
pile at a certain depth. The deflection at top of wall due
to a constant lateral loading increases accordingly byincreasing the effective wall height. This shows theeffect of piered retaining wall dimensions on the total
improve in the lateral capacity. The optimum effective
wall height can be determined using the numericalmodel developed in the present study.
A generalized improvement over a retaining wall can beshown by running this numerical model in Matlab. If thepercentage of load transferred to piles from the wall is low,
the piling impact is not substantial and using piered retaining
wall is arguable, since the use of piles beneath the retaining
wall will not considerably improve the total lateral capacity.
As the percentage of load transfer from wall to piles
increases, the impact of piles becomes more prevalent and thenecessity to utilize a piered retaining wall to achieve a desired
lateral capacity instead of a simple typical wall becomes more
obvious.
4 ACKNOWLEDGMENT
The author would like to acknowledge Dr. Wei Dong Guo forhis guidance, mentoring and support and Mr. Nick Thill for
his contribution during the numerical modeling process.
5 REFERENCES[1] Andraws, K. Z., and El-sohby, M. (1973). Factors
affecting coefficient of earth pressures K 0 . ASCE
Journal of Geotechnical Engineering, vol. 99.
[2] Caquot, A., and Kerisel, J. (1948). Tables for the
calculation of passive pressure, active pressure andbearing capacity of foundations. Paris: Gauthier- Villard.
[3] Kerisel, J., and Absi, E. (1990). Tables de pousse´e et debute´e des terres. 3rd ed. Presses de l’E´ cole Nationale
des Ponts et Chausse´es, Paris.
[4] Benmebarek, S., Khelifa, T., Benmebarek, N. and
Kastner, R. (2007). Numerical evaluation of 3D passive
earth pressure coefficients for retaining wall subjected totranslation. Computers and Geotechnics, 35, 47-60.
[5] Cundall, P. A. (1987). Distinct element models of rockand soil structure. Analytical and computational methods
in engineering rock mechanics. London: Allen & Unwin,
chap 4-pp187-190.
[6] Matlock, H., and Reese, L. C. (1960). Generalized
solutions of laterally loaded piles. Journal of
Geotechnical Engineering, ASCE, 86(5):63-91.
[7] Guo, W. D. (2003). A simplified approach for piles due
to soil movement. Proc. of the 12th Pan-American
Conference on Soil Mechanics and GeotechnicalEngineering, Cambridge, MIT, U.S.A. Vol. 2,2215-2220.
[8] Chen, Z. Y., and Li, S. M. (1998). Evaluation of active
earth pressure by the generalized method of slices.
Canadian Geotechnical Journal, 35(4):591-599.

654

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051


Evaluation of Thermal Response Tests in Energy Piles
Seung-Rae Lee1, Seok Yoon2, Hyun-Ku Park 3, Gi-Dae Oh 4
PhD. Professor, Dept. of Civil and Environmental Engineering, KAIST, Daejeon, Korea
2
PhD Student, Dept. of Civil and Environmental Engineering, KAIST, Daejeon, Korea
3
PhD. Post Doctor, Dept. of Civil and Environmental Engineering, KAIST, Daejeon, Korea
4MSc. Senior Researcher, Power System Lab, KEPRI, Daejeon, Korea
1

ABSTRACTThis paper describes an experimental and analytical study on thermal response test in precast-high strength concrete(PHC) energy piles. The thermal response tests (TRTs) of PHC energy pile were carried out using two type of groundheat exchangers (GHEs) : W and 3U-shaped heat exchangers. Ground thermal conductivity was measured by laboratory
test using transient hot wire method and inversely estimated from TRT results. Total thermal resistance was alsomeasured by TRT and compared with an analytical solution. Ground thermal conductivity of W-shaped heat exchanger
was higher, and total thermal resistance was smaller than 3U-shpaed heat exchanger because of different conditions
such as pile size and volume of grout.
Keywords: Ground thermal conductivity, Thermal Response Tests, Thermal Resistance, Ground Heat Exchangers

1. INTRODUCTION
Subsurface geothermal resources represent a great potentialof directly usable energy, especially in connection withfoundations and heat pumps. Geothermal energy used to becalled ubiquitous energy because it can be used at any time atany place. Geothermal energy can be obtained by means ofborehole heat exchangers. With geothermal cooling/heating
systems, heat energy is fed into and withdrawn from theground via GHEs. Conventional GHEs consist of U-shaped
GHEs inserted into a borehole. Moreover, recently, energy
foundations substitute boreholes in the heat exchanger system.
That is, GHEs can be embedded in piles, and thus it decreases
initial construction cost. Energy pile generally has shorter
length in the rage of several tens of meters. In Korean
practices, PHC piles are broadly used [1].
This research investigates an experimental and analytical
study of energy piles which are composed of 3U and W
shaped heat exchangers. PHC energy piles were installed at
an electric substation in Suwon, Korea. TRT results were
calculated using simple line source theory to derive effective
ground thermal conductivity. Thermal resistance was also
calculated and compared by both experimental and analyticalsolutions.
2. THEORETICAL BACKGROUND

mechanism of a ground heat exchanger is quite complex andconjugated because of various heat transfer mechanisms
involved inside and outside GHEs. Soil is a multiple phasesystem involving complex heat transfer mechanisms, but heattransfer in soils occurs mainly through conduction [2],whereby energy is passed from one region of a medium toanother by molecular transfer. According to Fourier's law, theheat flux through an arbitrary area (A) during time t can be
written as
.
Q
Q
T
(1)
q

 
At
A
x
where  is the thermal conductivity and ∂T/∂x is thetemperature gradient. Heat exchange rate of circulating waterin the GHE by conduction can be represented by temperature
difference between initial temperature of soils and fluid
divided by total thermal resistance. The heat transfer process
can be divided into steady state inside the borehole, andtransient state outside it. Thermal resistance inside theborehole is composed of fluid convection, pipe conduction,
grout conduction and pile conduction. Thermal resistanceoutside the borehole is that of conduction in the soil [3], [4].
[(ti  to ) / 2]  t
Q

L
R

(2)

Here, L: length (m)t : initial temperature of soils

2.1 Heat transfer mechanism of GHE
In the GHEs system, heat is extracted from or released tosurrounding ground through fluid circulating, for example, ina U-type pipe embedded in a borehole. The heat transfer

655

ti ,to : inlet and outlet temperature of GHEs, respectively
ΣR : total thermal resistance( R = Rsoil  Rpile  Rgrout  Rpipe  Rconvection )Based on equation (1) and (2), two kinds of models such as
analytical models and numerical models can be applied toobtain the thermal conductivity. The applicability depends on
the assumption and the methodology adopted in the models.
Among many models, a line source model is widely used forsimplicity and convenience. The governing equation of themodel represents a linear characteristic for long timemeasurement. The GHEs pipe can be assumed as a line source
because the radius ( rb ) is much smaller than the length ofborehole. An effective thermal conductivity (  ) can besolved by equation (3). In the equation, T f represents anaverage temperature of fluid that flows inside the borehole.
To is an initial temperature, Q is an input power (W), L is a2length of borehole (m), a is a thermal diffusivity ( m / s ),Rb is a thermal resistance of borehole, and  is an Euler
constant (0.5772).

Q/L
Q / L 4at
T f  To 
ln t ( ) 
4
4 rb 2
r2
Q
Q / L 4at
Rb  To ( 2 )(1  b )L4 rb
16at(3)

An average temperature of fluid ( T f ) can be expressed by a

1st linear equation with respect to ln t , and it can be

PHC piles were installed for an energy pile system (Fig.1).
Table 1 shows dimensions of energy piles and GHEs. 3U andW shaped polybutylene pipes (inner/outer diameter of
pipe=0.016/0.2m) were used. The depth and inner diameter of
PHC pile between 3U and W-shaped GHEs were not equal
because of construction condition. Cement grout with cement
to water ratio of 0.5 was poured and cured for more than 28
days. The ground was composed of weathered granite soil
and stiff weathered rock.
Ground water level was 4.5m below from the top of
embedded energy pile. Average void ratio e and water content
w of soil below and above the ground water table were
measured as follows: e=0.66, w=12.4% above G.W.T. and
e=0.45, w=16.7% below G.W.T.
Thermal conductivity of the weathered granite soil wasmeasured using TP-02 device (Hukseflux Thermal Sensors
Inc.), based on a transient hot-wire method. Soil was
remolded to have the same void ratio and water content of theconstruction site. Thermal conductivities of soils above andbelow ground water table were measured as 1.1 and 2.4
W/m·K, respectively.
Fig. 2 shows an experimental system for the thermal response
test. Constant heat flux was applied to water tank using anelectric heater, and TRTs were performed until steady state
condition was attained. Temperatures of water at inlet andoutlet were monitored during the tests, and flow rate was alsomeasured at outlet. Different heat flux and flow rate wereapplied for the energy pile with 3U and W-shaped GHEs asfollows: 1,000W and 17 lpm (liter per minute) for 3U case,
800W and 16 lpm for W case.

described by

T f  bx  c

(4)

4at
Q
Q/L
, x  ln t , and c  b( 2   )  Rb  To .
rb
L
4
Once b can be solved, an effective thermal conductivity (  )b

can be obtained by



Q/L
4 b

(5)

Therefore, fluid temperatures at inlet and outlet of GHEs canbe measured with respect to time by the thermal response test,
and an effective thermal conductivity can be obtained withthe value of the slope ( b ) from the T f  ln t relationship.
3. EXPERIMENTAL SETUP
3.1 Thermal Response Test

Fig. 1 Plane view of energy pile with 3U and W GHEs

656

Table 1 Dimensions of energy pile with 3U and W GHEs
Dimension of pile

3U

W

Pile Depth (m)

13.75

13.27

Pile Inner Diameter (mm)

240

225

Pile Outer Diameter (mm)

400

400

Data logger PC

sensors on pipes

Pump Heater
W
GHE
2
m

3U

Pipe installation

GHE
Fig. 2 Experimental system for thermal response tes
3.2 Test Results
The thermal response test was conducted for more than 30hours. Mean circulating fluid temperature-time curves
measured during TRTs are plotted in Fig. 3. Initial
temperature of soil was between 16~17 °C. Temperature of
fluid in 3U-shaped GHE reached to 30.19 °C, and 27.52 °C in
W-shaped GHE. As a mean circulating fluid temperaturereaches to lower value, heat can be well transferred to theground. An effective ground thermal conductivity can becalculated using Eq (5), and total thermal resistance can becalculated using Eq (2). The ground thermal conductivity of
3U and W-shaped GHEs was 2.14 and 2.32 W/m·K,
respectively. Total thermal resistances of 3U and W-shaped

657

GHEs were 0.1772 and 0.1702 m·K/W. W-shaped GHE has ahigher ground thermal conductivity and lower thermal
resistance than 3U-shaped GHE from experimental data.
Total resistances of 3U and W-shaped GHEs from theanalytical solution were 0.1998 and 0.1991 m·K/W. The
relative error between experimental and analytical results was
between 12~17%. W-shaped GHE has lower total thermal
resistance than 3U-shaped GHE. It is thought that there is adifference of inner diameter in PHC pile, and it came up withdifferent grout thermal resistance. Total thermal resistancefrom the analytical solution under the same condition withequal size of PHC pile and pile depth was also calculated.
Total thermal resistances of 3U and W-shaped GHEs above
the condition were 0.1953 and 0.1991 m·K/W, respectively.

32
30

Fluid Temperature (K)

28
26
24
22
20
18

W GHE
3U GHE

16
0

200

400

600

800

1000 1200 1400 1600 1800 2000

Elapsed Time (min)Fig. 3 Temperature variation of fluid
Table 2 Summary of TRTs
GHE
type

Input
Heater (W)Released
Heater (W)λgroundΣR, Experiement

ΣR, Analytical(W/m·K)(m·K/W)(m·K/W)

ΣR, Analytical(same condition)

3U

1000

997

2.14

0.17720.19980.1953

W

800

838

2.32

0.17020.19910.1991

It can be seen that 3U-shaped GHE has a little lower thermal
resistance than W-shaped GHE under the same condition.
However, it is necessary to consider a proper pile inner size to
consider heat transfer mechanism between 3U and W-shaped
GHEs.
4. CONCLUSION
An experiment system for thermal response test for energy
piles was established in the field. TRTs were performed fortwo prototype PHC piles with 3U and W-shaped GHEs in apartially saturated weathered granite soil and rock condition.
The ground thermal conductivity of 3U and W-shaped GHEs
was 2.14 and 2.32 W/m·K, and total thermal resistance of 3U
and W-shaped GHEs was 0.1772 and 0.1702 m·K/W from

658

TRTs, respectively. W-shaped GHE has a higher groundthermal conductivity and lower thermal resistance than
3U-shaped GHE based on experimental data. Total resistanceof 3U and W-shaped GHEs from analytical solution was
0.1998 and 0.1991 m·K/W, respectively. However, if thecondition is same, total thermal resistance of 3U and
W-shaped GHEs will be 0.1953 and 0.1991 m·K/W,
respectively. It can be concluded that 3U and W-shaped
GHEs show a very similar heat transfer mechanism for small
diameter energy piles. However, it is indispensible toconsider the effect like of inner size and pile depth in order toanalyze heat transfer in detail.
5. ACKNOWLEDGEMENT

This study was financially supported by National
Research Foundation of Korea funded by the Ministryof Education, Science, Technology (under grant No.
2010-00014639) and 2011 Construction Technology
Innovation Project, (11 Technology Innovation E04).
6. REFERENCES[1] Baek, S.K. An, H.J, Lim, S.K. Development of groundheat exchangers installed in building foundation.

659

Proceedings of SAREK Summer Annual Conference
2007. p. 277-282,[2] Brandl, H. Energy foundation and other thermo-active
ground structures. Geotechniques 2006, 56(2), p. 81-122.
[3] Liu Jun, Zhang Xu, Gao Jun, Yang Jie. Evaluation ofheat exchange rate of GHE in geothermal heat pumpsystem. Renewable Eenrgy 2009, 34, pp. 2898-2904[4] Min, H.S. Yun, T.S. Jeong, S.S. Effect of Group Spacingof Energy Piles on Thermal Analysis, Journal of Korean
Geotechnical Society 2011, 27(8), pp. 39-50.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Experimental Study on Mechanical Characteristics of Soil-Geosynthetic Interface
Md. Bellal Hossain, Md. Zakaria Hossain and Toshinori Sakai
Department of Environmental Science & Technology, Mie University, Japan

ABSTRACTThe shear deformation behaviour along soil–geosynthetic interfaces has been investigated and the results are comparedbased on the backfill material, types of inclusion, and interface mechanical properties. Three soils (sandy, clayey, andpure sand) in combination with four different geosynthetics (one geotextile and three geogrids) were tested at various
loading conditions in direct shear. Test results reveal that the stress-deformation behaviour of the geotextile and geogrid
interfaces with sandy and clayey backfills can be defined as hyperbolic. For the pure sand-geogrid interfaces, therelationship is followed by displacement hardening and softening behaviour. The dilatancy behaviour of a particularsoil-geosynthetic interface is found similar at all normal stresses. Both contractive and dilative nature is observed forthe interfaces with pure sand. On the contrary, only negative dilatancy or contractive behaviour is observed for sandyand clayey backfills with all geosynthetics. Based on the test results, a simplified equation has been proposed for the
nonlinear soil-geosynthetic interface shear strength envelops which was in good agreement with the experimental data.
Keywords: Geosynthetics, Interface behaviour, direct shear test, shear strength envelop

The practice of reinforcing the soil with various tensile
inclusions has been widely implemented in geotechnicalengineering for the last few decades. The interaction between
soil and reinforcement is of utmost importance for the design
and performance of reinforced soil structures, and thisinteraction can be very complex depending on the nature andproperties of the reinforcement. Various types of geosynthetic
materials have been used for soil reinforcement including
geotextiles (woven and nonwoven), geogrids and geocells.
Whatever the reinforcement and backfill materials are used
for the design of a reinforced soil structure, the interaction
properties of soil-reinforcement interface play an importantrole. The interaction mechanism between the reinforcement
and the soil can be classified into two types; sliding of soil
over the reinforcement and pullout of reinforcement from thesoil [1]. The direct shear and pullout tests are widely usedmethods to study quantitatively these interaction
mechanisms.
Practically, the direct shear test is a suitable mean to study theinteraction between soil and reinforcement because it cansimulate the shear mechanism along a potential failure plane
in reinforced earth structure. Proper representation of thesoil-geosynthetic interfaces under direct shear mode is also
important for numerical simulation of the deformation
response for retaining structures.
The shear strength of soil-geosynthetic interfaces has been
investigated using direct shear tests by a number of
researchers [1-9]. They used various types of geosynthetic
materials like High density polyethylene (HDPE), poly
propylene (PP), Polyester (PET) yarns coated with PVC,
basalt fiber geotextile and geogrids etc. The present study isthe continuation of the previous study conducted by the
Authors [9]. Here, in addition to Toyoura sand, two different
soils (sandy and clayey) have been used as backfill materials
to describe the interface behaviour for the same geosynthetic
materials. From test results, the shear strength parameters andstress-deformation behaviour have been compared anddiscussed briefly.
2. MATERIALS AND METHODS
2.1 Backfill materials
Three types of soils were used as backfill material in thisstudy. Air-dry pure silica sand named as Toyoura sand, one
sandy and one clayey soil have been used to evaluate theinterface behaviour with the same geosynthetic materials.
Before conducting the interface direct shear tests, the physical

660

100

Percent finer, %1. INTRODUCTION

80

Toyoura sandsandy soil
clayey soil

60
40
20
0
0.001

0.01

0.1

1

Particle size, mm
Fig. 1 Gradation curve of experimental soils

10properties of the soils were tested in the laboratory. The testsof physical properties reveal that Toyoura sand has no finecontent less than 0.075mm with effective diameter D 10 of0.01 mm, D 30 of 0.15 mm, D 60 of 0.19 mm, the uniformity
coefficient (Cu) of 15.45, and the coefficient of curvature of
the gradation curve (Cc) of 9.89. The specific gravity,
maximum and minimum void ratio are G s =2.64, e max = 0.98
and e min = 0.61, respectively. The water content of sand wasless than 1%, which corresponds to air-dried condition. The
optimum water content of sandy and clayey soils was 12%
and 15%, respectively. Fig.1 represents the gradation curvesof the experimental soils. According to JGS engineeringclassification system, the sandy and clayey soils are classified
as SF and CL, respectively. The major physical properties ofthe soils are listed in Table 1.

3.

TESTING METHODS

A small scale direct shear testing device which consists of a
fixed lower box and a moving upper shear box has been usedin this study. Both the shear boxes have same inside
dimension of 150mm in length and 100mm in width. Detailof this device is also available at [9]. The applied shear force,
horizontal and vertical displacements were recorded using
one load cell, and two displacement transducers (one forhorizontal and other for vertical displacement measurement)
connected to a computer through a data-acquisition system.
The geosynthetic specimens were positioned on a substrate

Table 1 Physical properties of the backfill soils
Properties

ClayeySandy
3Particle density, ρ𝑠

2.7g/cm

2.64g/ cm3

Water content, 𝑊𝑜𝑝𝑡

15%

12%

Liquid limit, LL

42.7%

---

Plastic limit, PL

25.0%

---

Plasticity index, PI

17.7

---

Sand (75µm-2mm)

41%

73%

Silt (5µm-75µm

32%

19%

Clay <5µm

20%

4%

GT

GG1

2.2 Geosynthetics
One woven geotextile and three geogrids, made of Basalt
fiber, are used in this study. These materials possess good
property of elasticity and stretch tension with good resistanceto acid, alkali, heat and vibration. It has also non-conductive
and non-magnetic resistance. The physical characteristics ofthese geosynthetics are listed in Table 2. For the purpose of
discussion, the geotextile is noted as GT and the geogrids arenoted as GG1, GG2 and GG3 as shown in Fig. 2.

GG2

Table 2 Physical properties of geosynthetic
Type of geosynthetics

Properties

GG1

GG2

GG3

GT

Specific weight, gm/m

165

165

350

450

Percent open area, %Machine direction

45
65
70
--Ultimate tensile strength(kN/m)
50
50
50
1870

Cross direction

40

2

40

50

GG3

1600

Fig. 2 Geosynthetic samples used in this study
661

662

GG2

GG3

GT

70

τ, kPa

60
50
40
30
20
10

0
0

1

2

3

4

u, mm

80
70

5

6

7

8

Backfill: Sandy soil

GG1

GG2

GG3

GT

60
50

τ, kPa4.1 Shear stress-displacement relationshipA series of direct shear tests were performed to obtain the
stress-deformation characteristics of the geosynthetic interfaces
for different backfill soils. The test results reveal that therelationship between shear stress and relative displacement
largely depends on the type of backfill rather than the type ofgeosynthetics. Fig.3 represents the stress-displacement
relationship between different backfill soil and geosynthetic
interfaces for the constant normal stresses of 40 and 160kPa.
Since, similar relationship is observed for the normal stressesof 80 and 120kPa, related data are not presented here. It isseen that the stress-displacement relationship of geotextile(GT) interfaces is same for all the backfill soils (Toyoura sand,sandy soil and clayey soil). The relationship for GT interfaces
can be defined as hyperbolic in nature where no displacementsoftening behaviour is observed. In case of pure sand,however, some insignificant softening behaviour is observed
at normal stress of 40kPa. In case of geogrid interfaces (GG1,
GG2 and GG3), with sandy and clayey soils, thestress-displacement relationship is found similar to its GTinterfaces. But, with Toyoura sand, the geogrid interfaces
produce a well defined peak and residual shear stresses for all
applied normal stresses. Before and after the peak shear stress,
displacement hardening and softening behaviour is clearlyobserved. Maximum shear resistance mobilizes within
3-6mm of shear displacement for geogrid interfaces with puresand backfill. The shear strength is found increasing with theincrease of normal stress which is a common phenomenon for
all soil-geosynthetic interfaces tested in this study.

GG1

80

40
30
20
10

0
0

1

2

3

4

u, mm

70
60

5

6

7

8

Backfill: Clayey soil

GG1

GG2

GT

GG3

50

τ, kPaRESULTS AND DISCUSSION

Backfill: Toyoura sand

90

σ=

4.

100

σ=

placed on the top of the lower box. Subsequently, thespecimen was clamped on the front edge of the substrate steel
plate using four aligned bolts and two steel clamping blocks.
In case of pure sand, the upper shear box was filled by raining
the sand from a height passing through two consecutive
sieves. The dry unit weight of the sand mass in upper box was
1.63 g/cm3 (15.9kN/m3) which corresponds to relative density(Dr) of 95%. The height of the sand layer in upper shear box
was 70mm. For sandy and clayey soils, water was added tothe soil samples to its optimum level before filling up theshear box. The upper box was filled by the soil in three steps
with same compaction energy for every step. Thus, thedensity of backfill soil was kept almost constant. The direct
shear tests were conducted using four different normal
stresses of 40, 80, 120 and 160 kPa. All the tests involved
applying the normal stress and monitoring the verticaldisplacement. The shear load was only applied after thevertical displacement had reached equilibrium. The normalload was maintained constant during shearing process. The
rate of shearing was maintained at 0.5mm/min for all the
tests. The maximum shear strength obtained during the shear
process was recorded as the peak shear strength. The sameprocedure was repeated for all types of geosynthetics. For
three soils and four types of geosynthetics, a total of 48 Nos.
of tests were conducted in this study.

40
30
20
10

0
0

1

2

3

4

u, mm

5

6

7

8

Fig. 3 Shear stress-displacement relationship of theinterfaces for constant normal stresses of 40kPa and
160kPa. τ, shear stress; u, relative displacement.

0.15

Backfill: Toyoura sand
GT

0.10

GG1

GG3

GG2

v, mm

0.05

0.00

160
-0.05

-0.10
0

1

2

3

4

u, mm

0.2

5

6

7

8

Backfill: Sandy soil
GT

0.0

GG2

GG1

GG3

4.3 Interface Shear strength
The shear strength of a soil-geosynthetic interface is an
essential parameter of slope stability analysis where slip
surface runs along the geosynthetic. The test results show thatthe interface shear strength values are not linearly dependent
on the normal stress. The relationship between interface shear
strength and normal stress at the interface is generally
considered to be linear and defined by (1).

v, mm

-0.2

-0.4

160
-0.6

-0.8
0

1

2

3

4

u, mm

0.2

5

6

7

8

Backfill: Clayey soil

0.0

GT

GG1

GG2

GG3

v, mm

-0.2
-0.4
-0.6
-0.8
-1.0

-1.2
0

1

4.2 Dilatancy behaviour
From test results, it is observed that the dilatancy behaviour ofa particular interface is similar for any normal stress. Thus ,for comparison, the vertical versus shear displacement curves
obtained from the direct shear tests on the soil-geosynthetic
interfaces at 160kPa applied normal stress are shown in Fig. 4.
Both contractive and dilative nature is observed for theinterfaces with the backfill material of pure sand. On thecontrary, for the sandy and clayey backfills, only contractive
nature is observed for all the interfaces (Fig. 4). For puresand-geosynthetic interfaces, the dilation with shearing
indicates the presence of some degree of particle rolling andinterlocking as dilation is required for the shearing andrearrangement of angular particles. For geogrid interfaces
with pure sand, it is interesting to note that the value ofmaximum vertical displacement reduces with the increase of
the percent open area of geogrid for a particular normal stress.
And, the amount of dilation is seen higher at lower normal
stresses and comparatively less at higher normal stresses.
However, no such relationship is observed in case of sandyand clayey backfills which were mostly fluctuating in nature.

2

3

4

u, mm

5

6

7

8

Fig. 4 Vertical vs. shear displacement relationships ofthe geosynthetic interfaces with different backfill
materials. v, vertical displacement; u, shear
displacement.

𝜏 = 𝑎 + 𝜎𝑛 𝑡𝑎𝑛𝛿(1)

Where, τ is the soil-Geosynthetic interface shear strength; 𝑎 is
the interface adhesion; 𝜎𝑛 is the stress normal to the interface;and δ is the interface friction angle. In many cases, resultsfrom shear box tests showed that the relationship between τand 𝜎𝑛 is not linear [10]. Usually, in those cases,
approximate linear relationships (1) are often considered.
This linear relationship may lead to significant errors where
factor of safety of a slope is concerned.
For utilizing interface shear strength data and to prevent theerrors, Giroud et al. [10] proposed a hyperbolic equation torepresent a nonlinear relationship consistent with interface
shear strength values obtained in laboratory shear box tests.
The equation contains five parameters to be determined, andthe determination of the order of hyperbola is a bit complex.
However, it is observed that the interface shear strength data
may be linear or nonlinear against the normal stressdepending on the characteristics of geosynthetic and backfill
materials. Therefore, it has been found that a similar type ofgeneralised nonlinear equation is appropriate to represent
both the linear and nonlinear peak shear strength data for thesoil-geosynthetic interfaces studied in this research. The
equation for interface shear strength, τ , can be expressed as

663

0

Where, 𝜏0 =
𝜎𝑛 =
𝛿0 =
𝑎0 =
𝑎𝑖 =
𝑚 =

𝑖

(2)

𝜎𝑛 𝑡𝑎𝑛𝛿0
Normal stress
Interface friction angle at very low
normal stresses.
Pseudo adhesion at 𝜎𝑛 = 0 (i.e. interceptof the nonlinear curve with τ axis).
Pseudo adhesion for higher values of σ𝑛 .
Curve fitting parameter; the value ranges
from 0 to 1.

If the experimental data is available for very low normal
stresses, 𝑎0 can be obtained by extrapolating the curve of τand 𝜎𝑛 . Otherwise, for practical purposes, 𝑎0 can be taken as
zero for the interface between geosynthetic and cohesionlesssoil. 𝑎0 may not be zero for the interface with cohesive soil ofhigh moisture content. In that case, the value of 𝑎0 can beobtained by best fitting the nonlinear curve with theexperimental data using a computer program. The interface
friction angle, 𝛿0 , is initially calculated using the data of very
low normal stresses. By the way of curve fitting, the value of𝛿0 can be optimized if the data of low normal stress is notavailable. The value of 𝑎𝑖 and m is completely depends on theavailable data of higher normal stresses. 𝑎𝑖 can be obtained
initially by fitting and extrapolating the straight line with theshear strength data points at higher normal stresses towards τaxis. The curve fitting parameter, m, is determined by themethod of least square which is easy by using a computer
program. Once m is obtained, the values of 𝑎0 , 𝛿0 , and 𝑎𝑖 canbe optimized, if necessary, to get higher coefficient ofregression (R2) value. In case of truly linear data, the value of𝑎0 and 𝑎𝑖 should be same and the value of m would be nearzero or may be negative. In that case, equation (1) isrecommended which is simple and convenient for linear data.
The proposed equation is not an empirical equation and theparameter values are adjustable based on the extent of theavailable experimental data. Thus, it may not be realistic to
use the equation for the purpose of prediction. Because it isfound that the value of 𝑎𝑖 and m is controlled by theexperimental data of higher normal stresses.
The equation is found suitable for better representation of
nonlinear experimental data and minimizing the errors whereappropriate shear strength value is very important such as inslope stability analysis of reinforced earth structure.
The shear strength envelops as shown in Fig. 5 weregenerated using (2) for the geotextile (Fig. 5a) and one
geogrid (Fig. 5b) interfaces with Toyoura sand (T), sandy soil(S) and clayey soil (C). Here, for pure sand, the value ofinterface adhesion, 𝑎0 , is taken as zero. The parameters valueof the shear strength envelops used in Fig. 5 are listed in Table
3. It is seen that the nonlinear shear strength envelops are ingood representation with the experimental data points.

CONCLUSIONS

The interface direct shear tests were carried out to evaluate
the mechanical properties of a new type of geosynthetic
material with three different backfill soils.
The constitutive behaviour of the geotextile and geogrid
interfaces with sandy and clayey soil is found similar innature which can be defined as hyperbolic relationship. In
case of pure sand-geotextile interface, the behaviour is also
hyperbolic. However, for sand-geogrid interfaces, theconstitutive behaviour is followed by displacement hardening
and softening behaviour. The dilatancy behaviour of aparticular soil-geosynthetic interface is found similar for all
normal stresses. Both contractive and dilative nature isobserved for the interfaces with pure sand. On the contrary,
only negative dilatancy or contractive behaviour is observed
for sandy and clayey backfills with all geosynthetics.
Based on the test results, a simplified nonlinear equation has
been proposed for the soil-geosynthetic interface shear
strength envelops which was in good agreement with theexperimental data.
100

Shear strength, kPa

𝜏 = 𝑎0 + 1 + 𝜏 𝑚 /𝑎

4.

T-GT
Nonlinear (T-GT)
S-GT
Nonlinear (S-GT)
C-GT
Nonlinear (C-GT)

80
60

a

40
20
0
0

50

100

150

200

Normal Stress, kPa
150

T-GG1
Nonlinear (T-GG1)
S-GG1
Nonlinear (S-GG1)
C-GG1
Nonlinear (C-GG1)

125
Shear strength, kPa

𝜏0

100
75

b

50
25
0
0

100

200

Normal Stress, kPa
Fig. 5 Nonlinear shear strength envelops for the geotextile (a),and geogrid (b) interfaces with different backfill soils.

664

Table 3 The parameter values of the interface shear strength
envelops plotted in Fig. 5Interfaces
T-GT
T-GG1
S-GT
S-GG1
C-GT
C-GG1
5.

𝑎0

0
0
1.5
0.5
3.1
8

𝑚
0.53
0.71
0.53
0.47
0.24
0.42

𝑎𝑖
10.5
31.2
10.0
15.0
7.9
10.0

𝛿0
37.7
48.5
24.2
23.9
15.5
16.3

R2
0.993
0.965
0.990
0.996
0.982
0.999

REFERENCES[1] Jewell, R.A. and Milligan G.W.E., Sarsby, R.W., andDubois, D., “Interaction between soil and geogrids”,
Polymer Grid Reinforcement, Thomas Telfold Limited,
London, March 1984, pp. 18-30.
[2] Ingold, TS., “Laboratory pull-out testing of grid
reinforcements in sand”, Geotech. Testing J.,Vol. 6, No.
3, 1983, pp. 101-111.
[3] Palmeria E M and Milligan GWE, “Scale and otherfactors affecting the results of pull-out tests of grids
buried in sand”, Géotechnique, vol.39, No. 3, 1989, pp.
511-524.
[4] Bergado, D. T., Chai, J.C., Abiera, H. O., Alfaro, M.C.,
Balasubramaniam, “Interaction between CohesiveFrictional Soil and Various Grid Reinforcements”,
Geotextiles and Geomembranes, Vol. 12, No. 4, 1993,
pp. 327 – 349.
[5] Bergado D. T., Shivashankar R., Alfaro M. C., Chai JC,
Balasubramaniam A. S., “Interaction behaviour of steel
grid reinforcements in a clayey sand”, Géotechnique,
1993, vol.43, No.4, pp. 589-603.
[6] Cancelli, A., Rimoldi, P., and Togni, S., “Frictional
characteristi cs of geogrids by means of direct shear andpullout tests”, Proc., Int. Symp. on Earth Reinforcement ,Kyushu Univ., Fukuoka, Japan, 1992, pp. 51–56.
[7] Cazzuffi, D., Picarelli, L., Ricciuti, A., and Rimold, P.,“Laboratory investigations on the shear strength ofgeogrid reinforced soils”, ASTM Spec. Tech. Publ.,
1993, 1190, pp. 119–137.
[8] Bakeer, R. M., Sayed, M., Cates, P., and Subramanian,R., “Pullout and shear test on geogrid reinforced
lightweight aggregate”, Geotextiles and Geomembranes,
vol. 16, No. 2, 1998, pp. 119–133.
[9] Hossain, M. B., Hossain, M. Z. and Sakai, T., “Interface
Behaviour of Basalt Geosynthetic with Sand Using
Direct Shear Device”, in Proc. 1st International
Conference on Geotechnique, Construction Materials
and Environment, Mie, Japan, Nov.21-23, 191-196,
2011.
[10] Giroud, J. P., Darrasse J., and Bachus, R. C., “Hyperbolic expression for soil-geosynthetic orgeosynthetic-geosynthetic interface shear strength”,
Geotextiles and Geomembranes, Vol. 12, No. 3, 1993,
pp. 275 -286.
665

Second International Conference on Geotechnique, Construction Materials and Environment
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

An Interpretation of Mechanical Properties of
Bentonite as a Non-linear Elastic Material
Y. TAKAYAMA, S. TSURUMI, A. IIZUKA, K. KAWAI
KOBE UNIVERSITY, JAPANS. OHNO
KAJIMA CORPORATION, JAPANABSTRACTDue to its low permeability and remarkable expansion characteristics, bentonite has an excellent potential to be used asa buffer in the disposal of nuclear waste. In order to examine the long-term safety of bentonite buffer in the disposal of
nuclear waste, constitutive modeling of bentonite materials and numerical simulations based on the constitutive models
have been actively pursued in recent years. Bentonite materials are usually modeled as a geo-material with elasto-plastic
isotropic compression/expansion and plastic dilatancy characteristics. However, based on recent experimental data, theauthors feel that mere constitutive modeling for geo-materials may not be applicable to bentonite materials. In thispaper, fundamental considerations of elasto-plastic constitutive modeling for fully saturated bentonite materials areinvestigated. According to uniaxial compression test data by Sasakura et al. (2003), where lateral earth pressure wasmeasured, it was found that the hysteresis response between loading and unloading processes does not appear. It
suggested that bentonite is an elastic material in which the swelling line corresponds to the normally consolidated line inthe ‘e-logp’ relationship. Also, according to triaxial CU test results of bentonite materials, it was seen that the effective
mean stress remains almost unchanged during shearing. This implies that bentonite does not have dilatancy
characteristics as defined by the critical state theory. Similar interpretation can be derived from the experimental results
obtained from a series of triaxial CD tests by Cui et al. (2006).

1. INTRODUCTON
Bentonite is known to be a material that has remarkable
expansion characteristics and very low permeability.
Therefore, bentonite as a material, is expected to be useful asa buffer in the geological disposal of nuclear waste facility.
As a result, its mechanical and mineralogical properties havebeen widely investigated. At nuclear disposal sites, bentonite
is expected to play a role as a buffer, preventing leakage ofradioactively polluted water by its abilities to self-seal and to
mitigate stresses generated due to creep of base rock. Also,
the geological disposal of nuclear material at such a facility
needs to be stable and safe for an extremely long period oftime. In order to address such problems, it is necessary to
predict the long-term behavior of geological disposal at anuclear waste facility and reliable numerical simulation
techniques are required that can reliably address these issues.
Constitutive modeling that describes the mechanicalproperties of a material is a key technique in such numericalsimulations. Constitutive models for bentonite materials havebeen formulated phenomenologically based on theelasto-plastic response under loading and unloading
processes and dilatancy characteristics during shearing, likeother clayey materials. However, in the case of bentonite
materials, we feel that the available experimental data to
support constitutive modeling is not sufficient. Remarkable
expansion characteristics on water absorption and very low
permeability of bentonite make it difficult to interpret its
fundamental mechanical behavior. The aim of this paper is toreconsider the fundamental mechanical behavior of bentonite

materials based on recent experimental data and to discuss thetheoretical formulation of constitutive modeling for bentonite
materials.
2. Isotropic compression and swelling characteristics ofsaturated bentonite materials
Usually, isotropic compression and swelling characteristics
are discussed based on the experimental data obtained fromtriaxial tests under a drained condition. However, in the caseof bentonite materials, available experimental data on thecompression and swelling characteristics obtained fromtriaxial tests under the drained condition are neither sufficient
nor reliable. The primary data used to justify constitutive
models have come from oedometer tests. Compression andswelling behavior of bentonite materials obtained from theoedometer tests are not directly applicable for discussing
compression and swelling characteristics under isotropic
confining pressure, because the lateral stress cannot bemeasured in the oedometer tests. If the lateral stress could
reliably be estimated, more intensive discussion and morestrict interpretation on the compression and swellingcharacteristics of bentonite materials would have beenpossible. In this paper, we cite experimental data on bentonite
materials where lateral stress was measured with a special
oedometer test apparatus by Sasakura et al. (2003). We also
discuss the compression and swelling characteristics ofbentonite materials under isotropic confining pressure, which
are the fundamental mechanical properties for constitutive
modeling.

666

10
5

1
0.5
1

where K 0 − NC is the coefficient of earth pressure at rest under
the normal compression process, K 0 − OC is the coefficient ofearth pressure at rest under the unloaded swelling process,
OCR is the over-consolidation ratio defined by σ v′0 σ v′ ,σ v′0 is the pre-consolidation vertical stress, σ v′ is the current
vertical stress and m is the evaluation constant. According to
Fig. 2, Eq. (1) with m =0.6, it seems to satisfactorily explain

K0

1
0.9
0.8
0.7
0.6
0.5
4

5

10
OCR

50

100

Fig. 2 Relationship between K0 value and OCR（2）Isotropic deformation characteristics
Fig. 3 shows one dimensional compression (consolidation)
characteristics of Kunigel V1, obtained from oedometer tests
by Sasakura et al. (2002) and Ishikawa et al. (1997).
Normally, the consolidation lines by Sasakura et al. (2002)
and Ishikawa et al. (1997) do not correspond to each other.
This is because the nature of the bentonite slightly differs
depending on the year it was produced and the content of themontmorillonite could be slightly different even in the same
bentonite sample.
According to Kobayashi et al. (2007), one-dimensional
normal consolidation characteristics of bentonite under fully
0.8

3
2

Void ratio, e

Horizontal effective stress,
σ'r MPa

Void ratio, e

2

m=0.4
m=0.5
m=0.6
Test result

K0（1）K 0 compression and swelling test
Sasakura et al. (2003) carried out one-dimensional
compression and swelling tests for Kunigel V-1, which istypically made of sodium bentonite and is planned to be
used in nuclear waste disposal sites in Japan. The authors
tried to measure lateral stress under one-dimensionalloading and unloading processes with a special oedometer
test apparatus. Their experimental results are shown in Fig.
1. It was found that the coefficient of earth pressure at rest,
K 0 , remains almost at 1.0 under the normal compression
process and then increases with OCR under the unloaded
swelling process. The relationship between K 0 value andOCR is indicated in Fig. 2. In this figure, experimental data
by Sasakura et al. (2003) is indicated by white circles and iscompared to the empirical relation, Eq. (1) which relates
K 0 -value and OCR through the evaluation parameter m as:
K 0 − NC = K 0 − OC OCR m （1）the experimental results of bentonite, Kunigel V-1. If tone
employs Eq. (1) with 0.6 as the value of m , in converting thevertical stress into the isotropic confining pressure,fundamental mechanical properties on compression andswelling characteristics under isotropic confining pressure
can be discussed using experimental data obtained from theoedometer test.

1
10
4

Ishikawa et. al ,1997
Ishikawa et. al ,1997
Sasakura et. al ,2002
Fully saturation curve

0.6

0.4

2
1

0.01
0.1
1
10
100
Vertical effective stress, σ'v (MPa)0.010.050.1
0.5 1
Vertical effective stress, σ'v (MPa)Fig. 3 One-dimensional compression (consolidation)
characteristics of Kunigel V1Fig. 1 K0 compression and swelling test result
667

saturated state is uniquely expressed as:

e

λρ ln

σ v′ maxσ v′

σ v′ max  Gs − λρ ln ′ σv σ v′ max = 3.73GPa .
（2）

The effective vertical stress can be converted into aneffective mean stress using Eq. (1) as:
σ ′ + 2 K 0σ v′（3）p′ = v

where λρ is the material constant, Gs is the specific gravity ofsoil particle and σ v′ max is the reference and maximum verticalstress corresponding to the void ratio vanishing. This curve iscalled the full saturation curve. In the case of data by
Sasakura et al. (2002), λ ρ = 0.156 , G s = 2.733 ,Void ratio, e

0.7

Sasakura et.al, 2002
Fully saturation curve

0.6

0.5
0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)

0.4

3. Dilatancy characteristics of saturated bentonite
materials
In this section, we examine the dilatancy
characteristics of bentonite. Usually, dilatancy
characteristics are examined by conducting triaxial
CD tests in which the principal mean effective stressis maintained as a constant. However, in the case of
bentonite materials, there is not sufficient reliabledata being obtained from such tests. Therefore, we
examine the dilatancy characteristic using theresults of the triaxial CU test on bentonite.

0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)（1）Triaxial CU test
Results of triaxial CU test on a few kinds of bentonite
are shown in Fig. 5-8. These figures show the
relationship between effective mean stress and

(A)

0.6

Void ratio, e3where p ′ is the effective mean stress, σ v′ is the effective
vertical stress and K 0 is the coefficient of earth pressure at
rest. All data in Fig. 3 has been converted in terms of theeffective mean stress using Eq. (3) and is depicted in Fig. 4 asa relation between the void ratio and the effective mean stress,
in which K 0 − NC ＝ １ and ｍ ＝ 0.6. It is found that thehysteresis-loop, between loading and unloading processes,
disappears and the swelling line corresponds to thecompression line in Fig. 4. If the compression and theswelling responses in Fig. 4 can be interpreted as isotropic
compression and swelling characteristics of bentonite, thebentonite can be regarded as a material with no plastic
deformation. In other words, it can be said that the bentonite
is a non-linear elastic material.

Ishikawa et. al, 1997
Fully saturation curve

Deviator stress , q (MPa)(B)

Void ratio, e0.8
Ishikawa et. al, 1997
Fully saturation curve

0.6
0.4
0.1
0.51
5 10
50 100Effective mean stress, p' (MPa)

3
2
1

0
0 1 2 3 4 5 6
Effective mean stress ,p' (MPa)(C)
Fig. 4 Isotropic compression and swelling characteristics
of saturated bentonite materials

Fig. 5 Triaxial CU test result on sodium type ofbentonite conducted by Sasakura et al. (2002)
668

Deviator stress , q (MPa)

Deviator stress , q (MPa)

5
4
3
2
1

0
0 1 2 3 4 5 6
Effective mean stress ,p' (MPa)

4
3
2
1

0
0
5
10
Effective mean stress ,p' (MPa)Fig. 6 Triaxial CU test result on calcium type ofbentonite conducted by Sasakura et al. (2002, 2003)

Deviator stress , q MPa

5Fig. 8 Triaxial CU test result on MX-80 conductedby Ann et al. (2010)
materials suck water to keep uniformity in the volume.
Typically, these results are interpreted as a connection
between pressure at saturation to dry density. However, therelationship between the results of this test and full saturationline are not perfectly correlated. Saturated bentonite is anelastic body, so it always lies on the full saturation line.
Because of this, we estimate that the final value of theseswelling tests reach the full saturation line. Below, we
examine the relationship between swelling characteristic andfull saturation line, based on the results of this swelling test.

2

1

0
0
1
2
3
Effective mean stress ,p' (MPa)Fig. 7 Triaxial CU test result on OT-9607
conducted by Takaji et al. (1999)
deviator stress. According to these test results, it canbe seen that the effective mean stress is almost
unchanged during shearing. Usually, the effective
mean stress decreases with increasing shear stressbecause of negative dilatancy, when triaxial CU test
on normally consolidated clay is conducted. It istherefore implied that the bentonite does not havedilatancy characteristics based on the critical state
theory.
4. Swelling characteristic of saturated Bentonite
An important mechanical property of bentonite is theswelling characteristic associated with the process of
saturation from an unsaturated state. There are two kinds oftests for measuring swelling characteristics: the swellingdeformation test and the swelling pressure test. Swellingdeformation test reports deformations when the bentonite
materials suck water to control the stress, and the swellingpressure test measures the swelling pressure when bentonite(1) Bentonite’s swelling deformation test
Cui (2006) conducted swelling deformation tests andisotropic unloading tests by controlling principal stress.
These tests were performed on samples of compacted
Kunigel V-1/Toyourasa mixture with Kunigel V-1 content of
50% in dry mass. In the tests, they set the initial dry density atthree patterns: 1.69, 1.54 and 1.2g/m3; and then performed
water absorption tests under hydrostatic stress, based on thecondition that the tests maintain mean principal stress settled.
The sample’s initial values, initial dry density and initialwater content ratio, and the value of mean effective principal
stress under which the samples suck water in are shown in
Table-1.
Additional isotropic unloading tests were also performed,
these reduce the load under 49kPa after the sample suckedwater with mean principal stress set at 392 kPa. Fig. 9 showsthese results: the relationship between mean effective
principal stress and void ratio at the finish of retting andswelling of these three cases of initial dry density as well as
the relationship between mean principal stress and void ratio
when the test goes on isotropic unloading. In the case that it
gave the sample some water in same mean stress, Fig. 9shows that the void ratio takes the same value regardless of
initial dry density. Furthermore, the value of unloading passat isotropic unloading shows almost the same result as theswelling deformation test. This leads us to interpret that threewords have the same meaning: full saturated line, normal
consolidation line, and isotropic unloading line. Based onthese results, one can conclude that the value, in the case of669

the swelling test, lies on the full-saturated line under saturated
conditions. Next, swelling tests were conducted with different
stress ratios R( = σ a / σ r ) where the principal mean stress andthe principal stress ratios were held constant. The initialvalues of the sample are: 1.69g/m3 for dry density and 18.7%
as the initial water content ratio. We show the relationshipbetween the void ratio and the mean principal stress at the endof soaking in Fig. 10. This figure shows that the void ratio andthe mean stress exhibit a one-to-one correspondence
regardless of the principal stress ratio. So, our examination
ρd0（g/cm3）18.7
49,98,147,196
18.7 19.6,49,88,196,294,392
19
49,98,196,392
Table 1 Initial valuethat suggests saturated bentonite is a nonlinear elastic
material without the characteristic of dilatancy coincides withthe result of Cui’s test.

Void ratio , e105ρd0=1.69g/cm3
ρd0=1.54g/cm3
ρd0=1.20g/cm3
Isotropic unloading
392→49kPa

1
0.5

50 100500 1000

Effective mean stress , p' (kPa)void ratio , eFig. 9 Result of amount of swelling test

Void ratio, e

Fully saturation curve
Suzuki et.al, 1999
Komine et.al, 2002

0.5
0.1

p
（kPa）

w0
(%)

1.69
1.54
1.2

0.1
10

1

0.51
5 10
50 100Swelling pressure (MPa)Fig. 11 the result of swelling pressure test(2) Swelling pressure test for bentonite
A swelling pressure test, using equipment similar to that of
the one-dimensional consolidation equipment, is a test thatmeasures the counter force along the vertical direction caused
by vertical soaking under settled volume. Fig. 11 shows theresult of the swelling pressure test of Kunigel V-1, and therelationship between the void ratio and the vertical swellingpressure at saturation. This figure conveys that when the void
ratio is smaller, the swelling pressure becomes larger and theswelling pressure value varies in the same void ratio, but thevalue never crosses the full-saturated line. However,considering the results obtained so far, these separate
variability values result from measuring only verticalswelling pressure. So, if we take account of lateral stress andreconsider it using mean stress, we consider that all values areon the fully saturated line.
5. Conclusion
This study examines the mechanical characteristics ofbentonite based on recent experimental data. The conclusions
are summarized as follows:10.5(1) The swelling line at S r = 1.0 corresponds to theNormal Consolidation line. We also show thatsaturated bentonite does not have dilatancy
characteristics. Therefore, saturated bentonite isa non-linear elastic material.
(2) When water is supplied and the unsaturated
state is transformed to a saturated state, thefinal, or saturation, points achieve the fullsaturation curve.

0.1
10

In Fig. 12, the schematics of relationship between void ratio
and the principal mean stress during loading, unloading andswelling are summarized.

10
5

R=2.0
R=1.5
R=1.0

50

100

500

Effective mean stress ,p' (kPa)
Fig. 10 Results of swelling test670

eFully saturation curveor
NCL or Swelling line
at Sr=1.0

InitialSwelling
Pressure Test

Swelling
Deformation
Test

Initialln pFig. 12 Schematics of relationship between void ratio
and the principal mean stress

REFERENCES[1] SASAKURA, T., KUROYANAGI, M., KOBAYSHI, I.,
OKAMOTO, M. “Studies on mechanical behavior of
bentonite for development of the constitutive model II”,
2003, JNC TJ8400 2003-048. (in Japanese)[2] SASAKURA, T., KUROYANAGI, M., OKAMOTO, M.
“Studies on mechanical behavior of bentonite fordevelopment of the constitutive model”, 2002, JNC
TJ8400-2002-025. (in Japanese)[3] ISHIKAWA, H., ISHIGURO, K., NAMIKAWA, T.,
SUGANO, T. “Consolidation Properties of Buffer
Material” 1997, PNC TN8410 97-051. (in Japanese)[4] KOBAYSHI, I., TOIDA, M., SASAKURA, T. and
OHTA, H. “Interpretation of compression/swelling
behavior of compacted bentonite using constantwater-content line and constant degree-of saturation line”,
2007, JSCE Journal of Geotechnical Engineering, 63(4):1065-1078. (in Japanese)[5] TAKAJI, K., SUZUKI, H. “Static Mechanical Properties
of Buffer Material” 1999, JNC TN8400 99-041. (inJapanese)[6] Ann Dueck, Lennart Börgesson, Lars-Erik Johannesson.
“Stress-strain relation of bentonite at undrained shear”
2010, SKB Technical Report TR-10-32.
[7] Hongbin CUI, De’an SUN, MATSUOKA, H. “Swellingcharacteristics of Sand-Bentonite mixtures underisotropic and anisotropic stress States” 2006, JSCE
Journal of Geotechnical Engineering, 62(3): 657-666. (inJapanese)[8] SUZUKI, H. and FUJITA, T. “Swelling Characteristics
of Buffer Material” 1999, JNC TN8400 99-038. (inJapanese)[9] KOMINE, H. and OGATA, N. “Swelling characteristics
of sand-bentonite mixture and various kinds ofbentonite”, 2002, JSCE Journal of GeotechnicalEngineering, 701(III-58): 373-385. (in Japanese)

671

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effect of carbonation on strength of cement treated sand1

Kenichiro NAKARAI1 and Tomomi YOSHIDA2
Hiroshima University, Graduate School of Engineering, Japan
2
Gunma University, Graduate School of Engineering, Japan2

ABSTRACT
It is well known that the strength of cement treated soil increases for several years. The long-term increase in thestrength is thought to be caused by pozzolanic reaction between cement paste and soil. On the other hand, carbonationof cementitious material can increase its strength. In this study, the effect of carbonation of cement treated sand on its
strength was investigated. In the experiment, the specimens of the cement treated Toyoura sand were cured in three
conditions; sealed condition, natural condition, accelerated carbonation condition. The concentrations of CO2 in the
curing conditions are 0%, 0.03%, and 5%. The measured strength of the specimens cured in the sealed condition didn’t
increase after 28 days while the strength of the specimens cured in the air or accelerated carbonation condition increased
after 28 days due to carbonation. These results indicated that the carbonation can be one of the main reasons of thelong-term increase in the strength of the cement treated sand.
Keywords: Cement, carbonation, strength, sand

1. INTRODUCTION
Cement treated soils, which are the soils mixed with cements,
are widely used to improve the properties of foundations. It is
well known that the strength of cement treated soil increases
for several years. The long-term strength after a few years
becomes several times of the strength at 28 days [1], [2]. It is
unique property of the cement treated soil because theincrease in the strength of concrete after 28 days is usuallyless than twice. The mechanism of the increase in the strengthof the cement treated soils usually explained by the
combination of the original strength of the soil without
cement, improvement of the basic soil properties due to thedecrease in water contents, cement hydration and pozzolanic
reaction (Fig. 1 [3]). Continuous long-term increase in thestrength of the cement treated soils is explained by the
pozzolanic reaction between soils and cement paste due tohigh alkaline condition. However, there is no quantified

Strength ofcement treated soil
Pozzolanic reaction
Cement hydration

Improvement of mechanical property
Original strength of soil

investigation on the contribution of the pozzolanic reaction.
On the other hand, the cementitious material such as cement
treated soil and concrete are affected by carbonation.
Carbonation is a process by which carbon dioxide in theambient air penetrates the cementitious materials and reacts
with the calcium hydroxides and other cement hydrated
products in the cement paste to form carbonates. A lot of
studies on carbonation of concrete have been publishedbecause the carbonation is one of the main reasons of thecorrosion of steels embedded in reinforced concrete
structures. According to the past studies in the concrete
engineering, the carbonation significantly lowers thealkalinity by the consumption of calcium hydroxides while it
makes pore structure dense by the precipitation of carbonate.
The decrease in the alkalinity causes the steel corrosion while
the densification increases in the strength [4]. Because thespeed of the carbonation of the porous cementitious materials
such as cemented treated soil is faster than that of normal
concretes, the influence of the carbonation on the cement
treated soils may be larger.
In this study, the effect of carbonation of the cement treatedsand on its strength was investigated. In the experiment, thespecimens of the cement treated Toyoura sand were cured inthree conditions; sealed condition, air condition, accelerated
carbonation condition. Then, the strength development andthe amount of carbonates were analyzed.
2. OUTLINE OF EXPERIMENT

Age

Fig.1 Proposed mechanism of increase in strength of cement
treated soils [3]672

2.1 Specimens
The cement treated specimens were prepared by mixing
Toyoura sand, ordinary Portland cement and water. The
designed values of the cement to sand ratio, the water content,
the water to cement ratio, the void ratio, the density of thespecimen were 8 %, 6 %, 70 %, 30 % and 2.8 Mg/m3,

respectively. After mixing, the cylindrical specimen having a
diameter in 100 mm and a height in 200 mm were compacted
by a weight of 2.5 kg. The number of the compaction times
was 25 for one of three layers of the specimen.
Table 1 shows the curing conditions of the specimens. The
specimens for investigating the effect of the carbonation were
cured in the sealed condition for 5 days and then cured in thewater for 1 day. After that, the specimens were cured in theenvironmental test room (natural condition) and chamber(accelerated condition). The average concentration of CO2,
the temperature, and the relative humidity in the natural
condition were around 0.03 %, 20 °C and 60 %, respectively
while the concentration of CO2, the temperature, and therelative humidity in the accelerated condition was controlled
to be 5 %, 20 °C and 60 %, respectively, to promote
carbonation of the specimens. In addition, the specimens
cured in the sealed condition were prepared to investigate thestrength development excluding carbonation effect.
Table 1 Curing conditions of the specimens(T=20°C)

Method of curing [days]
1-6
6-7
7Room
Water
Room
(T=20°C,
RH=60%,
CO2=0.03%)(T=20°C)

Accelerated(CO2=5%)
Sealed(CO2=0%)3.1 Results of unconfined compressive strength
Figs. 2 show the measured stress-strain relation of therepresentative specimens cured in the natural and accelerated
conditions. Table 2 shows the compressive strengths and theelastic coefficient of the specimens. Each data is the averagevalue of three results under the same condition. Figs. 3 andFig 4 show the changes in the compressive strengths and theelastic coefficient with time, respectively.
In the sealed condition, the strength of the specimen wasalmost constant after 28 days until half a year. In the natural
conditions, the strength of specimen constantly increased.
The strength at the age of 180 days was 1.4 times of thestrength at the age of 28 days. In the accelerated carbonating
condition, the strength initially increased and then decreased.
The maximum strength at the age of 56 days was 1.3 times of(T=20°C,
RH=60%,
CO2≅0.03%)

6

Chamber(T=20°C,
RH=60%,
CO2=5%)

Sealed (T=20°C)

2.2 Unconfined uniaxial compressive test
After prescribed curing periods, the unconfined compressive
strength of specimens were investigated by the uniaxial
compressive tests based on JIS A1216. For filling the gap
between the specimen and the loading plate of the test
machine, the gypsum was used. The displacement of thespecimen cured in the natural or accelerated condition wasmeasured by using LDT.
2.3 Thermo-gravimetric analysis
The amounts of calcium hydroxide (Ca(OH)2) and calcium
carbonate (CaCO3) were measured by using thethermo-gravimetric analysis (TGA). The sample was takenfrom the part of around 50 mm from the bottom of thecylindrical specimen after the compressive test and it was
grinded. Then, the changes in the mass of the powder sample
were measured from room temperature to 1000 °C. The
temperature rise rate was set to 10 °C/min. The amount of
Ca(OH)2 was determined by the mass decrease between 390
and 450 °C while the amount of CaCO3 was determined by
the mass decrease between 450 and 750 °C. In the discussion,
they were modified based on the measured chemically boundwater between 105 and 1000 °C to consider the variety of thecomposition of cement paste and sand in the sample.

673

5

Stress [N/mm2]

0-1
Sealed3. RESULTS AND DISCUSSIONS

4
3
28days56days91days180days

2
1
0
0

0.1

0.2

0.3

Axial strain [%](a) Specimens cured in the natural condition (CO2≅0.03%)
6
5

Stress [N/mm2]

Curingcondition
Natural(CO2≅
0.03%)

2.4 X-ray diffraction analysis
Crystalline phases, especially phase of CaCO3 (calcite,
vatelite, and aragonite), were identified by X-ray diffraction(XRD) analysis for the power sample after the compressive
test.

4
3
28days35days42days56days91days

2
1
0
0

0.1

0.2

0.3

Axial strain [%](b) Specimens cured in the accelerated condition (CO2=5%)
Figs. 2 Strain-stress relation

Table 2 Measured unconfined compressive strength

Initial curing(1day sealed, 5days dried, 1day
water, and then 21 days dried)
Natural(CO2=0.05%)

Accelerated(CO2=5%)

Sealed(CO2=0%)Age[days]
28

Strength[N/mm2]
3.69

12

E50
[kN/mm2]
3.71

56

4.42

5.21

6

91

4.42

5.13

5.12

5.85

35

4.10

6.16

42

4.46

5.38

56

4.79

5.17

91

4.60

9.80

180

4.20

Fig.4 Changes in elastic coefficient due to carbonation

180

3.29

4.08
-

365

3.48

-

3.22

-

the strength at the age of 28 days.
The elastic coefficient, E50, of the specimens showed almost
similar behaviors as the strength. In the natural conditions, theelastic coefficient constantly increased. On the other hand, in
the accelerated carbonating condition, the elastic coefficient
initially increased and then decreased. The maximum elastic
coefficient in the accelerated condition was observed at theage of 91 days, which is later than the age for the maximum
strength.

1052

5Strength 「 N/mm2]

8

180

6

4

4
Natural2

Accelerated

0
0

50

100

150

200

250

Age [days]

3
Natural2

Accelerated
1

Sealed

0
0

250

500

750

1000

1250

Age [days](a) Data in whole periods
6
5Strength 「 N/mm2]

10E50 「 kN/mm2]

Curing condition

4
3
Natural2

Accelerated
1

Sealed

0
0

50

100

150

200

250

Age [days](b) Data until half a year
Figs.3 Changes in strength under different curing conditions

674

3.2 Results of thermo-gravimetric analysis
Figs 5 and 6 show the results of the thermo-gravimetricanalysis on Ca(OH)2 and CaCO3, respectively. Here, we
calculated the ratios of Ca(OH)2 and CaCO3 to cement. In thecalculation, the amount of chemically bound water was
assumed to be 23% of the cement considering the past study[5].
Regardless to the CO2 concentration in the atmosphere, theamount of Ca(OH)2 decreased and the amount of CaCO3
increased due to carbonation with time. The changes in theaccelerated condition were larger than that in the natural
condition because the high CO2 concentration accelerates the
progress of the carbonation.
Fig. 7 shows the relationship between the increase ratios of
the unconfined compressive strength and the amount of
CaCO3. Here, the increase ratios of the strength and CaCO3
after 28 days were calculated. In the natural condition, thestrength was almost constantly increased with the increase in
CaCO3. In the accelerated condition, the strength was firstly
increased until 56 days and then decreased with the increasein CaCO3. The increase ratio of the strength to CaCO3 (theslope of plots in Fig. 7) in the natural condition was larger
than that in the accelerated condition. These results suggested
that the effect of the carbonation on the strength depends onthe concentration of CO2 and the carbonation progressed in
the low CO2 concentration effectively increases the strength.

analysis were indicated.
At the age of 56 days, the main phase of the generated CaCO3
was analyzed as calcite. On the other hand, the main phase of
the generated CaCO3 at the age of 91 days in the accelerated
carbonating condition was analyzed as vaterite. It suggested
the carbonation of C-S-H gel and the transformation of calcite
to vaterite in the accelerated condition after 56 days. By
comparing with the results of the strength in Fig. 1, it wassuggested that the generation of vaterite caused the decreasein the strength in the accelerated condition.

35
30
NaturalCa(OH)2 [%]25
20

Accelerated

15
10
5
0
0

50

100

150

200

250

Age [days]

Fig.5 Changes in amount of Ca(OH)2 under different curingconditions

Table 3 Peak strength measured by X-ray diffraction analysis
CuringAgecalcitevateritearagonite
condition[days]Natural56+++++(CO2=0.05%)

Accelerated

56
+++
+
+
91
++
+++
+
+++: strong peak, ++: weak peak, + very weak peak, blank:
no peak(CO2=5%)

60

CaCO3 [%]

50
40

4 CONCLUSION

30This study investigated the effect of carbonation on thestrength development of the cement treated Toyoura sand byusing the specimens cured in three conditions; sealedcondition (CO2= 0%), natural condition (CO2≅0.03%),accelerated carbonating condition (CO2=5%). Then, theexperimental results showed the followings.
In the sealed condition, the strength was almost constant after
28 days. The carbonation in the natural condition constantly
increased the strength and elastic coefficient. The carbonationin the accelerated condition firstly increased the strength andelastic coefficient and then decreased them. The results of
XRD analysis suggested that the generation of vaterite caused
the decrease in the strength in the accelerated condition.

20

Natural10

Accelerated

0
0

50

100

150

200

250

Age [days]

Fig.6 Changes in amount of CaCO3 under different curingconditions

Strength / Strength_28days

1.5
180days

1.4

5 REFERENCES

56days1.3

91days91days56days1.2
1.1

180daysNatural1
28days0.9

Accelerated

0.8
0

1

2

3

4

5

CaCO3 / CaCO3_28days

Fig.7 Relationship between unconfined compressive strengthand CaCO3 (increase ratios after 28 days)
3.2 Results of X-ray diffraction analysis
Table 3 shows the results of the X-ray diffraction analysis.
Here, the peak strengths of the phases of CaCO3 (calcite,
vatelite, and aragonite) measured by X-ray diffraction675[1] Horpibulsuk S., Miura N., & Nagaraj T.S., “Assessmentof strength development in cement-admixed high watercontent clays with Abrams' law as a basis,”
Geotechnique, 53 (4), 2003, pp. 439–444[2] Kongsukprasert L., Tatsuoka F., & Takahashi H.,“Effects of curing period and stress conditions on thestrengthanddeformation
characteristics
ofcement-mixed soil,” Soil and foundation, 47(3), 2007,
pp.577-596.
[3] Express highway research foundation of japan,“Research report on planning, design, and constructionof cement/lime stabilized roadbed,” 1978, p.6. (inJapanese)[4] Leber I. and Blakey F.A., “Some effects of carbon
dioxide on mortars and concrete,” Journal of theAmerican Concrete Institute, 28(3), 1956, pp. 295-308.
[5] Powers T.C., “The Physical structure of Portland cement
paste,” The Chemistry of Cement, edited by H. F. Taylor,
Academic Press, 1964, pp.391-416.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Application of PML to Analysis of Nonlinear Soil-Structure-Fluid

Problem Using Mixed Element

Pahaiti Reheman1, Hiroo Shiojiri2,
1
Graduated School of College of Science and Technology, Nihon University, Japan
2
College of Science and Technology, Nihon University, Japan
SUMMERY:
Mixed element may be conveniently used to express non-linear constitutive equation of fluid and to avoid volumetriclocking. X-FEM may be well suited to model discontinuity of displacements between solid and fluid. Appropriate
boundary conditions should be set at the boundaries of numerical models not to reflect outgoing waves. In this paper,complex frequency shifted convolution-PML without splitting of variables is developed for mixed element, and theperformances of PML are confirmed. The formulation of PML is completely consistent with corresponding FEM or
X-FEM. It can be easily extended to any type of element and any nonlinear constitutive equations of the corresponding
FEM or X-FEM. The resulting mass and stiffness matrices for PML are symmetric for linear models.
Keywords: Soil-structure-fluid interaction, X-FEM, Mixed formulation, Complex Frequency shifted PML

1. INTRODUCTION
Soil-structure-fluid interaction may have significant effects
on seismic responses of structures. Mixed element may be
conveniently used to express non-linear constitutive equationof fluid and to avoid volumetric locking. X-FEM may be well
suited to model discontinuity of displacements between solidand fluid. In the X-FEM analysis, as well as FEM and FDM
analyses, appropriate boundary conditions should be set at theboundaries of numerical models not to reflect outgoingwaves.
Several methods are proposed (Wolf 1988). The first is theextensive mesh models using a finite element method or afinite difference method with approximate energy
transmitting boundaries. The second is the substructure
method using, for example, finite element and time domainboundary element method. In the former, the degrees of
freedom of the models are often very large. The latter method
may be more efficient, but the nonlinearity must be restrictedwithin the nearby portion of structures modeled by finite
element method, i.e. ,constitutive equations are assumed to belinear at and outer domain of the boundary. The third is FEM
with PML or convoltion PML(Berenger 1994, Collino
2001,Basu 2003 2004,Drossaert 2007). PML and convolution
PML are proved to have efficient wave absorbing capability
for linear elasto-dynamic problem, and, the nonlinearity must
be restricted within finite element domain. In the severe
earthquakes, however, soil may become nonlinear to a large
extent so that the second and the third methods may beinadequate. Convolutional PML is extended to cope with
non-linear problem, so that nonlinear soil can be analyzed

677with a limited number of meshes without loss of accuracy(Shiojiri 2010, Reheman 2011). But, it is restricted todisplacement based FEM.
Here, complex frequency shifted convolution-PML without
splitting of variables is developed for mixed finite element
and for X-FEM, and the performances of PML are confirmed.
The formulation of PML is completely consistent withcorresponding FEM or X-FEM. It can be easily extended to
any type of element and any nonlinear constitutive equationsof the corresponding FEM or X-FEM. The resulting mass andstiffness matrices for PML are symmetric for linear models.
2. METHOD
2.1 PML Formulation of Mixed Element for Fluid
Assuming that the effect of viscosity is negligible and thatchange of density is small, the equations of motion of fluid isgiven as:

 0ui 


p0
xi

(2.1)

,where 0 is time averaged density of fluid, p is dynamic
pressure, ui is with component of displacement, and xi
denotes i th coordinate. Relationship between density anddisplacements is given as:

ui
0
i xi

  0 (2.2),where  is difference between current density and time
averaged decity. Relationship between dynamic pressure and is expressed as follows.

p  f ( ),where f (  ) is given as:(2.3)

f (  )  c 2  for p  pv  p0
f (  )  (  0 c )   (1   0 )( pv  p0 )22for p  pv  p0

 q  p  f (  ) dv  0(2.3a)

Considering the fact that  , i ,and k1   1 /(1  i )2(2.3b),in which c is velocity of sound, pv is vapor pressure, p0 is
static pressure ,and  0 is reduction ratio of sound velocity

in frequency domain corresponds to dxi

written as:v

Equations of motion are given as:


p0
xi

S

e

 t

(2.7)

sides of Eqn.2.6 and Eqn.2.7 by 12 ,we get;
(2.8)

ui 
0
 xi 

(2.9)

,where i denotes an integer other than i. Introducing weightfunctions wi for displacements, and q for pressure, weakiform equations for Eqn.2.8 and Eqn.2.9 are given as follows.

  wi 12 2  0ui dv  VV   wi ni  i pds* ui  dvwi
 i pdvxi(2.10)

S

677(2.12)

k j  j ( j   j )   1 2

 j  j* f (t )  0 e

 ( t t )

e

 jt

*

(2.13)

Denoting

f (t ) dt   F (t ) ,and introducing approximation

F (t  t )   t (1   ) e   t f (t )   f (t   t )  e   t F (t ) ,we get  w r u ( t   t ) dv
0

v

i 12

i

2

v

Considering  / xi  (1/ i ) / xi , and multiplying both j t   0 w i  U ij* ( t )dv,where  is relative density amplitudes in frequency domain.

1 2    0    itdisplacement and pressure amplitudes in frequency domain
respectively. The relationship between density anddisplacement are given as:

p
0
xi

e

 ui
u 
  0   k i
  i e  i t * i x xi 
i i,where * denotes convolution integral.

,where  is circular frequency, and u and p are 12 2  0ui   i

 j  j 1 2   

(2.6)

u
  0  i  0
ii x

k j  j ( j   j )   1 2

ij

, where i is pure imaginary number, ω circular frequency,
and k i , α i and σ i non-negative continuous functions, suchthat k i =1, and σ i =0 at FEM-PML interface. At first, allequations are formulated in x i coordinate in frequency
domain, and then transformed to x i coordinate.

  2  0ui 

1 2i   wi ni k i p   i e   i t * p ds

(2.5)

α i +iω0transformed coordinate, and λ i is given as:

λ i＝k i +/ dt 2 , d / dt ,and

j

 wi
k i p   i e   i t * p dv

V x
i

, where xi denotes i th coordinate, and x i the correspondingi  w  k k u  

(2.4)

0

2

k   ei t * in time domain, equations in time domain areafter cavitation.
Following the PML procedure, we introduce complex
coordinate stretching function in frequency domain as:x i =  λ i (s)ds(2.11)

V

V(2.14)j 1

 wi
r i p ( t   t )  Pi* ( t ) dv
 xi   w i n i r i p ( t   t )  Pi* ( t ) ds
S,wherer12  k1k2  t (k21  k1 2 ), ri  k i  t i ,and we

define as followings:t
Pi* ( t )   i  e   i  t  e   i ( t  t  ) p ( t ) dt 
0





 e   i  t  t (1   ) p ( t )  e   i  t Pi* ( t   t )   t i p ( t )(2.15)

k j  j ( j   j )   1 2

U ij* (t ) 

e

 r N
V

 j  j

 j tt

e j ( t  t  )

0

e

t



p

V

T

p

12

p

V

T

12

*


ui (t ) e   i  t [ R *i ( t   t )   t { k i  i ( i   i )   1 2 }/ { ( i   i ) c 2 } P e ( t )]Let H and G be defined asu   0  ri i (t  t )  V12 (t )xiiH   r12 N p T N p dv , G   N p T R Tm B dv . Then,we get:V

(2.17)

 ( t   t ) 0  r1 2 N s N s d v u  0 c 2 G T H  1G u e ( t   t )   N sT N R m N p d s P e (t   t )Si  N sT N S d s P (t )ie  i t [ Ri* (t  t )  t  (t )[k i  i ( i  i )V

Discritizing domain of analysis by finite element, adopting

N s and N p denote

displacement and pressure vector at nodal points of a element,
and interpolation matrices for displacement andpressure ,respectively, matrix form equations are obtained.
From Eqn.2.14, we get, ( t   t )   0  N s U ( t ) d v
 0  r1 2 N s N s d v uTvV(2.19)

  B S d v P (t )TV   N sT N R m N p d s P e (t   t )
S

T

1

mode and may lead to unstable solution. Since vorticity ispreserved in non-viscous flow, and since fluid is assumed to
be static in the beginning, the following equation is valid:uuc  2  1=0(2.22)
x1 x2
So, in PML domain, the following equation is assumed.
uu12 c  2 2  1 1
(2.23)
x1
x2


2 c  1 c  0(2.24)
x1
x2,where  c is amplitude of vorticity in frequency domain.

Converting Eqn.2.23 into time domain ,and introducing
approximation for convolution integral, we get,2 k  (   )   
1 2
j
j
j
j
r12 c (t  t ) E *j (t ) j  jj 1(2.25)


u j
*
j 1

(t  t )  U jj (t ) ( 1) r j
x j

j 






  N sT N S d s P (t )S,where

 2

 

*
0 
  U 1 j (t )  xn j 1
1
 N s , N   1
U (t )   2
,B   0
U 2* j ( t ) 
 0
 

 x 2 
j 1

0 
N p
 P * (t ) , P ( t )   2*
R Tm   r2 r1  , S  ,Np P1 ( t )  0  B T S d v P (t )2  qc 2  R12 (t )  V12 (t ) dv  B T R m N p d v P e (t   t )V

Element stiffness matrix  0 c G H G has many hourglass



u j
2





q
r
p
t
t
c
r
t
t
(
)

(
)

 dv 

j
12
0
V 
(2.18)
x j
j



e

1

T

V

From Eqn.2.17, Eqn.2.3a,and Eqn.2.11,we get,e  0  N s U (t ) d v  G H R (t )T,

Vi* (t )  e i t Vi* (t  t )  t 0 i ui / xi e(2.21)

S

1 2 }/( i  i )]Galerkin’s formulation, and let u , P

e

V

*

12

V

T

 R (t ), V (t )  V (t ) ,and Ri* (t )  ,v

 N N dvR (t )   N V (t ) dv 、

R 12 (t )  R 1* (t )  R *2 (t ) ,and R i (t ) (2.16)
 j  j(2.20)

Vr12  (t  t )  R12 (t )T

N p d v P e (t   t )R (t )   c 2

k j  j ( j   j )   1 2*
iT

,whereLikewise, from Eqn.2.13,we get:,where R12 (t ) p  0 c 2  N p T R Tm B d vu e ( t   t )  R ( t ) tui (t )dt   e j t (1   )ui (t ) *
U ij (t  t ) j t12,where0 ,
n 2 

, (t  t )  t (t )
U (t )  e
U (t  t )  t u (t ) / x  , andU (t )  e
U (t  t )  tt u (t ) / x  .
*j

 j t

*
12

1t

 (t )  e*
21

 2 t

*
j

c

*
12

1

*
21

1

2

2

2

2

Introducing weight function w1 , w2 and deriving weak form

P i* ( t )  e   i  t P i* ( t   t )   t  i P e ( t ) equation from Eqn.2.24,we get,From Eqn.2.18, we get,

678



 
    w2  2 c  w1 1 c  dv
V
x

x2 
1


(2.26)     w2 n1  2  w1 n 2 1   c ds
S  w2w 
    2
 1 1   c dv  0
V x1
 x2 


Transforming Eqn.2.26 into time domain equation, andintroducing approximation for convolution integral, andsubstitutingEqn.2.25,we get,w  u2u
  wr2(t  t )  r1 1 (t  t )  dv
V r12  r2 x12  r1 x21 


x
x
1
2


 w

w
    2  2 E2*  1 1E1*  dv
V
x2
 x1


(2.27)  w2



V



w 
 r1 1  E(t )dv
 r2
x2 
r12  x1

Fig.3.1. Reservor Mesh(length 100m)    w2n1r2  w1n2r1   c  w2n1 2 E2*  w1n21E1* ds
Swall on the left end of reservoir is assumed. The results areshown in Fig.3.4~3.9. Fixed boundary results show strong
dependency on the location of upstream boundary. The
results with viscous boundary show less dependency on thelocation of boundary for 1Hz excitation, but large
dependency for 5Hz excitation. Proposed PML boundaryshows very little discrepancy between different boundary
locations for both 1Hz and 5Hz excitation.

Fig.3.2. Reservor Mesh(length 300m),where E (t ) 2

{k  (   )    }E (t) / (   )  U (t) U (t) ,



j

j

j

1 2

j

j 1

*
j

j

*
21

j

*
12

 t
*j (t )  e j  (t t) t (t) ,U (t)  e  U (t t) t u (t)/ x  ,and U (t )  e   U (t  t )  t t u (t ) / x  .
*j 2 t

*
21

 1t

*
12

c

*
21

*
12

1

2

1

2

2

Fig.3.3. Reservor Mesh(length 600m)

1

Introducing interpolation matrices, the following equations
are obtained.

 r B B dv u (t   t )
V

c

T

e

c

12

   BeVT E1* 
B c T E ( t ) dv
 *  dv  V
r12
 E2 

(2.28),where




0 
 1

x2


 
 Ns
B c    r1
r2
 N s , B e  
 
x1 
 x2
2
 0

x1 

By adding Eqn.2.21 and Eqn.2.28,we get final matrix form
equations. Substituting interpolation function for X-FEM, we
get PML for XFEM.

Fig.3.4. Fixed Boundary (1Hz)

3 NUMERICAL EXAMPLE
3.1 Reservoir Model
Reservoir with 100m depth subjected to horizontal rigid wallmotion is analyzed using proposed mixed formulation
FE-PLM. Length of reservoir model varies from 100m to
600m. Boundary conditions assigned on upstream boundary
are fixed boundary, viscous boundary with viscosity  c , andPML. Meshes of models are shown in Fig.3.1~3.3. Mesheswith sign x indicate PML. Forced horizontal vibration of rigid

679

Fig.3.5. Fixed Boundary (5Hz)

Fig3.6.Viscous boundary(5Hz)

Fig 3.10.Pressure Distribution(1Hz)
3.2 Dam- Reservoir-Foundation Model
Dam-reservoir-foundation model was constructed using
X-FEM for tangential discontinuity of displacement at
solid-fluid interface ,and combining proposed mixed
formulation FE-PML for fluid with FE-PML for solid(Fig3.11) 。Two kinds of mashes are used (Fig.3.12,Fig.3.13).
Discontinuous displacement in tangential direction near the
top of dam is shown in Fig.14, indicating the effect of
discontinuous interpolation function of X-FEM. In Fig.15-17,
responses of dam top using two mesh models subjected tosinusoidal horizontal ground motion are compared. The
results of two mesh model coincide each other, indicating thatmodel size has little effect when proposed PML is used.

Fig3.7.Viscous boundary(5Hz)

10m

Reservoir Length(100m or 300m)

PML(10meshes)
100m
110m

80m
PML (10meshes)Fig 3.8. PML(1Hz)

PML (10meshes)Fig 3.11. Dam-reservoir-foundation model

Fig3.9.PML(5Hz)

Fig 3.12. Meshes (reservoir length=100m)

680

210m

Fig 3.13. Meshes(reservoir length=300m)

Fig 3.17. Response of dam top (5Hz)
4 CONCLUSION
Convolution PML based on mixed formulation is formulated
and computer code is developed for FEM and X-FEM. It wasapplied to reservoir model and the performance of absorbing
outgoing wave was much better than conventional boundary.
By applying to dam-reservoir-foundation model including
FEM and X-FEM, it was demonstrated that the formulation
was quite general, and that it has wide class of application.
5 REFERENCES

Fig 3.14. Displacements near top of dam

Fig 3.15. Response of dam top (1Hz)[1] Wolf, J.P.(1988). Soil-Structure-Interaction Analysis intime Domain, Prentice-Hall, Engelwood Cliffs.
[2] Berenger P.(1994).A Perfectly Matched Layer for theAbsorption of Electromagnetic Waves, Journal ofComputational Physics, 114, 185-200..
[3] Collino,F. ,and Tsogka,C.(2001).Application of theperfectly matched absorbing layer to the linear
elastodynamic problem in anisotropic heterogeneous
media, Geophysics, 66 :1,294-307.
[4] Drossaert,F.H.,abd Giannopoulis,A.(2007) Complex
frequency shifted convolution PMl for FTDT modelingof elastic waves, Wave Motion ,l.44:8593-604.
[5] Basu,U ,and Chopra, A.K.(2004),.Perfectly Matchedlayers transient elestodynamics of unbounded domains,
Int.J.Numerical Methods Eng..59, 1039-1074.
[6] Basu,U. ,and Chopra, A.K.(2003), Perfectly matchedlayers for time-harmonic elesto dynamics of unboundeddomains: theory and finite-element implementation,
Comput. Methods Appl. Mech. Eng.192, 1337-1375.
[7] H.Shiojiri,,P.Reheman. & Li J.F.(2010) Application of
PML for Non-Linear Soil-Structure Problem,.
14ECEE,Ohrid, 1-8CD.
[8] P. Reheman,H.SHIOJIRI, Jeng-Feng LI(2011)
Analytical Method for Nonlinear Soil-StructureInteraction using PML,Eurodyne, Leuven, Belgium,
1-6CD.

Fig 3.16. Response of dam top (3Hz)681

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Recycled Bassanite as a Stabilizing Agent for Cohesion-less Soil
Aly Ahmed1 and Keizo Ugai2
Department of Civil Engineering, Beni-Suef University, Beni-Suef, Egypt
1, 2
Department of Civil and Environmental Engineering, Gunma University, Gunma, Japan
1

ABSTRACTThis study evaluates the use of recycled bassanite, which is derived from gypsum waste plasterboard, to enhance theperformance of two types of cohesion-less soil. Recycled bassanite was utilized as a stabilizing agent to improve bothcompressive and splitting strengths of tested soil. The effect of bassanite content, soil type, water content, and curing
time were investigated to explore the behavior of treated soil with recycled bassanite. Test results showed that increaseof bassanite content is associated with increase in optimal moisture content, while no significant increase in the dry unit
weight was observed. Both compressive and splitting tensile strengths enhanced with the additives of recycled
bassanite. The increase of bassanite content had a more significant effect on the compressive strength compared withthe effect on tensile strength. The use of recycled bassanite to enhance the strength of sandy soil had a more significant
effect compared with silty soil. The effect of curing time on the strength of treated samples was more significant in earlycuring ages compared with late curing ages. The strength decreased significantly in case of stabilized samples preparedwith water content at the wet-side of the compaction curve. However, insignificant decrease in the strength of thestabilized sample was detected with moisture content at the dry-side of compaction curve.
Keywords: Soil Stabilization, Waste Materials, Soil Strength, Recycled Bassanite

1. INTRODUCTION
Nowadays, researchers are focused on obtaining a desirable
way to eliminate waste materials by using them asalternatives in the applications of civil and industrial
engineering. Gypsum waste plasterboard is considered one
example of these waste materials. Large quantities of gypsum
plasterboard, approximately 80 million tons in the world, areproduced annually. Around 85% of this production accounts
for Europe, USA, and Japan. Out of the latter figure, about 15
million tons is sent to landfill sites [1]. Plasterboard is madefrom gypsum sheets covered on both sides with paper sheets.
Subsequently, increasing the amount of gypsum waste
plasterboard results in more quantities, which are disposed inlandfills. Hence, it is deemed necessary to find alternative
solutions to such problems due to the following reasons. First,
gypsum waste plasterboard in landfills is known under certain
circumstances to cause hydrogen sulfide gasses, which arepotentially harmful and/or lethal. Second, there is currently
an increase in the cost of disposal in landfill and decrease inavailable ground spaces specified for landfills sites. Third,
landfill regulations in Japan require that quantities based on
gypsum materials are disposed to landfills, and that the waste
must be sent to control waste disposal site, which in turn leads
to increase in disposal cost.
Several researchers have conducted investigations to improve
the physical and mechanical properties of soil by using
different types of waste materials, such as cement kiln dust,
fly ash, bottom ash, blast furnace slag, stone dust, andfactory-waste roof shingles [2-6]. Most of the previous

studies investigated the use of recycled gypsum waste
plasterboard in agriculture applications, while a few studies
focused on the use of recycled gypsum/bassanite in civil andconstruction engineering applications. The new unfired
gypsum-clay blocks were manufactured by using processed
plasterboard from recycled sources. The blend of production
blocks comprises 50% of recycled plasterboard. Results
indicated that the product has a good physical properties
compared with natural clay [7]. Laboratory and field
investigations were conducted to develop effective novel
cementation mixes using recycled gypsum and a range ofmineral wastes for road foundation construction. For thispurpose, a series of laboratory trials were carried out todetermine the optimum proportions of recycled gypsum andmineral wastes of a binder paste to achieve the highest
compressive strength. The optimum combination for thenovel binder was a mix of 15% recycled gypsum from waste
plasterboard, 5% bypass dust, and 80% basic oxygen slag,
which gave a high compressive strength. Results indicatedthat plasterboard waste can be used in low-strength concrete
mixes for the foundations of minor roads and car parks [8].
Generally, recycled bassanite derived from gypsum waste
plasterboard has a potential to be used as a cementationmaterial because gypsum is the main key of cementation
products. Therefore, the main objective of this research is toinvestigate the use of recycled bassanite derived fromgypsum waste plasterboard as a stabilized agent to possibly
improve the strength of cohesion-less soil.

682

2. MATERIALS AND METHODS
Three types of materials, which include two different types ofsoil, recycled bassanite, and Portland cement, were used in
the current study. The two soil types used were brought from
Gunma prefecture, Japan. The two types of soil were named
as soil-A and soil-B. All physical and mechanical propertiesof tested soils are shown in Table I. The grain size distribution
curves for all tested soil are shown in Figure 1.
Recycled bassanite used in this work was derived fromgypsum waste plasterboard. The crushed plaster gypsum
waste (CaSO 4 .2H 2 O) was heated at a temperature ranging
from 130 to 160°C for a certain time results in a hemi-hydrate
calcium sulphate (CaSO 4 .0.5H 2 O) with three quarters of thewater removed. Firstly, gypsum waste plasterboard waspulverized and then screened to remove any impurities suchas papers, paints, and any solid material. Afterward, thepowder of gypsum waste plasterboard was placed on
specified plate and then the process of heating was started.
The preparation of recycled bassanite was done in someconstruction companies in Gunma prefecture, Japan and
Figure 2 is a schematic diagram for preparing the recycled
bassanite. Four different contents of recycled bassanite 5%,10%, 15%, and 20% per soil weight were investigated.
The ordinary Portland cement was used in this research with
content of 3% by dry soil mass, which was added to the tested
soil sample. The main reason for adding the cement was toprevent the solubility of gypsum when water was introduced
since hydrate calcium sulphate (CaSO 4 .2H 2 O) is a known
water-soluble material.

Figure1. Grain size distribution curve for used soils

Figure 2. Schematic diagram for producing recycled
bassanite from gypsum waste plasterboard

Table 1: Physical and mechanical properties for used soils
Property
Effective diameter, (D 10 ), (mm)Nominal mean size, (D 50 ), mm)
Uniformity coefficient, (C u )Curvature coefficient, (C c )Passing from sieve No. 200
Max. dry unit weight, (N/mm3 )Optimum moisture content,(%)
Specific gravity, Gs
Max. void ratio
Min. void ratio
Soil classification (USCS)
Soil classification (AASHTO)

Soil A
0.15
0.21
1.53
0.94
0.40
1.57e-5
13.20
2.60
0.85
0.60
SP
A-3
Fine sandSoil B
0.06
0.15
2.72
1.02
11.30
1.43 e-5
20.40
2.67
1.28
0.79
SP
A-2-4
Silty
sand

3. SAMPLE PREPERATIONThe effect of using recycled bassanite on the strength oftested soil was evaluated by conducting a series of unconfinedcompression and splitting tensile strength tests. Cylindricalsamples having a 50-mm diameter and 100-mm height wereused. The cylindrical mold, which was used to form the soil
samples, was made from high grade plastic with 5-mm
thickness to resist any external pressure during the process ofcompaction. The reason of using plastic molds was to preventthe friction between soil samples and the inner sides of themold during sample extraction. In addition, the use of oil tolubricate the inner sides of the mold was to make sure that nofriction will be occurred during sample extraction. The
samples were placed into the molds and compacted staticallyusing a static method by hydraulic Jack to reach the required
density. By knowing mold volume, which is constant for all
tests and the desired density, the weight of soil was
determined to validate the required density. Homogeneity ofthe samples was of main concern during preparation andextraction. The soil samples were extracted from the molds
after 24 hours. More details for preparing stabilized soil
samples were described previously [6]. All samples tests wereconducted after seven days of curing at temperature (21 ± 1°C) and relative humidity > 90%. Some samples were tested
after 3, 14, 28, and 56 days of curing under the same
conditions to investigate the effect of curing time on thestrength of samples treated with recycled bassanite.
All tests were conducted at the optimal moisture content andmaximum dry unit weight to simulate field conditions. A
standard compaction test was carried out to determine the dry
density/moisture-content relationship of untreated and treatedsoil with different contents of recycled bassanite, according to
ASTM D 698-70 specifications. The oven-dried soil andrecycled bassanite mixtures was firstly dry mixed then theamount of water tap was added to obtain the moisture content
for all tested soils.

683

4. RESULTS AND ANALYSIS
The dry density/moisture-content relations for soils A and B
treated with different contents of recycled bassanite areshown in Figures 3 to 4, respectively. It is clear that theincrease of recycled bassanite content is associated with asignificant increase in optimum moisture content in both twosoil types used. It was attributed to two reasons: firstly, due tothe tendency of hemi-hydrate calcium sulphate(CaSO4.0.5H2O) to obtain the three quarters of the water thatwas missed before during heating process, in order to change
to hydrate calcium sulphate (CaSO4.2H2O). The second
reason is related to when fine content such as recycled
bassanite was mixed with soil, more water was required incompaction in order to achieve the maximum dry unit weight[9]. In the case of soil type B (silty soil), the increase of
bassanite content up to 15% was associated with a slightly
increase in dry unit weight. This was attributed to the calcium
component in bassanite, which promotes soil particles toflocculate. Thus, attraction between soil particles wasdeveloped due to this property of flocculation. It may be
assumed that the flocculation occurred due to the finecomponents of the soil which normally carry negative
charges and attract calcium, which carries two positive
charges. This behavior could be related to the tested soil-B
comprising more than 10% content of silty clay, which cansubsequently behave as fine minerals. The reason of theslightly increase of dry unit weight is related to the increase of
optimal moisture content in the soil-bassanite mixture ascompared to pure soil. Consequently, the maximum dry unit
weight increased slightly. While in the case of soil type A(fine sand), slight increase in dry unit weight was related tothe particles of bassanite which are very fine compared to soil
particles. That makes it easy for some bassanite particles tofill the small pores within the particles of sandy soil. It isdifficult, from a practical view, to fill the pores completely
during compaction process in the case of pure sand. For thesame volume of compacted sample, the weight of thebassanite-sand mixture was slightly higher than the weight ofpure sand that was associated with the slight increase in dry
unit weight. On the contrary, if the difference between soil
particles was high such as in coarse sand, the dry unit weight
would be decreased.

Figure 3. Dry density/moisture-content curves for tested
soil-A at different contents of recycled bassanite used.

Figure 4. Dry density/moisture-content curves for tested
soil-B at different contents of recycled bassanite used.
The results presented herein are only for soil-B because thesame results were obtained with another type of soil-A. The
results of unconfined compression tests, stress-strain
relationship, for soil-B type treated with different contents ofrecycled bassanite are shown in Figure 5. Obviously, thecompressive strength increased with the increase of bassanite
content. These results suggested that recycled bassanite canbe used as stabilizer material in ground improvement. These
results are in agreement with the previous results [10, 11].
Improvement indexes for both compressive and splitting
tensile strengths against the content of bassanite are shown in
Figure 5. Improvement stress index was obtained by dividing
the ultimate stress for treated sample to the ultimate stress ofidentical un-treated sample. Figure 6 indicates that bothcompressive and splitting strengths were enhanced with theincrease of recycled bassanite content. The improvement incompressive strength was more significant than that in tensile
strength. This is attributed to stabilized soil cement, which is
weak in tension compared with compression state. An
increase of bassanite content up to 15% enhanced theimprovement of splitting strength compared to the otherlower contents used, as shown in Figure 5. To sum up, the useof bassanite as a stabilizing agent to improve the strength ofcohesion-less soil can be acceptable in both compressive andsplitting strengths.
Figure 7 shows the stress-strain relationship for different twosoil types treated with 10% content of recycled bassanite. It
can be observed that the higher performance was obtained in
the case of soil-A compared with the case of soil-B. The useof bassanite had a significant effect on the strength
enhancement of sandy soil compared to silty soil. This effectcould be due to the friction between soil particles of sandy
soil was higher than that in silty soil. Besides, the developed
cohesion between particles of sandy soil was greater than thatin silty soil due to the contact surface area between sand soil
particles and bassanite is strong. Thus, the effect of using
bassanite to enhance the strength of sandy soil was muchpronounced compared to silty sand soil. The same behavior
was obtained in the case of splitting tensile strength.

684

Figure 5. Stress-strain relationships for soil-B with different
bassanite contents

Figure6. Improvement index for both compressive andsplitting tensile strengths in case of soil-B treated withdifferent bassanite contents

Figure 7. Stress-strain relationships for different soil types
used treated with 10% content of recycled Bassanite.
To investigate the effect of water content condition on thestrength of soil treated with recycled bassanite, three different
moisture contents including optimal moisture content (OMC)

and moisture contents in dry and wet sides that corresponding
to 95% of maximum dry unit weight, were used, respectively.
The main reason for the use of 95% maximum dry unit weight
was to simulate filed conditions. Unconfined compression
test was conducted on treated samples of soil-B with recycled
bassanite contents of 5% and 15%, respectively for different
investigated cases of water contents. The suggested contents
of 5% and 15% of recycled bassanite were representing thesmallest and highest content in this study. The tests wereconducted after 7 and 14 days of curing age to investigate theeffect of water content on the age of treated samples. The
effects of water content on the strength of treated samples areshown in Figures 8 and 9. It is clear that for both age ofsamples, the high strength was obtained with samples
prepared at optimal moisture content compared to other two
used water contents. These results are consistent with the
expected routine results of compaction test. The effect ofwater content on the strength of 7-day old samples was muchpronounced than the 14-day old samples. The latter difference
may be due to the increase of curing age of the samples. This
was associated with the increase of strength and consuming
most of the excess water as a result of the reaction between
soil particles and bassanite. Also, it can be noted that theeffect of water content on the samples having highest
bassanite content was more prominent than the samples
having smallest bassanite content. This behavior could be
related to both increase in bassanite content and watercontent, as illustrated before. Subsequently, the effect ofwater content in the case of the samples having the highest
bassanite content was significant especially at moisture
content in the wet side. Furthermore, these figures in the caseof samples having smallest bassanite content of 5% showed
not much difference between the strength of samples
prepared at moisture content in dry side and samples preparedat optimal moisture content. This result also did not matchwith the theory of compaction test because these samples
were impure soil. However, they were treated with somecementation materials thus their behaviors were different. In
other words, the amount of moisture content in this casemeets the required amount of water to complete a goodreaction between cementation material and soil particles.
Additionally, the decrease of water content was associatedwith the increase of strength. Generally, it was found that theunconfined compressive strength decreased significantly inthe case of water content at wet side that corresponding to
95% of maximum dry unit weight, while the strength of watercontent at dry side was slightly decreased.
The effect of curing time on the compressive strength of
untreated and treated samples with the smallest content of 5%and largest content of 15% was investigated. Consequently,
five different curing times included 3, 7, 14, 28 and 56 days
were studied. As aforementioned, all samples were subjected
to curing under temperature (21 ± 1 °C) and relative humidity
more than 90%. Figure 10 shows the effect of curing time on
the compressive strength for tested soil-B at different curing
times. The compressive strength for all cases, treated and685

untreated samples, increased as curing time increased. Thisfinding is consistent with the theory of curing time forcementation materials. The proportion of improvement in thestrength of treated samples in the early curing days was
significantly higher than later curing time days. In otherwords, the improvement in strength for treated samples after
seven days curing time was increased slowly in relation with
the increase of curing time. This behavior could be explainedby the short time required to complete the setting ofsoil-bassanite mixture, subsequently the treated samples
regained most of their strength. Hence, the improvement instrength for treated samples was evident especially in the firstdays of curing after which the rate of improvement was
sluggish mainly with the largest gypsum content. It is worth
mentioning that after 14 days of curing time for bothbassanite contents, the effect of curing time was insignificant
in the context of the strength improvement. The early strengthfor soil-bassanite mixture is actually a vital property in someprojects, such as embankments. This property helps to open
embankments for trafficability after compaction process
directly. Subsequently, the time of construction decreases andthat leads to a reduction in construction cost.

Figure 8. Ultimate strength against water content for soil
type-B treated with bassanite and tested after 7 days curing.

Figure 10. Ultimate strengths for soil type-B tested at
different curing times.
5. CONCLUSIONSThis research meets the demand to reduce the quantities ofgypsum waste plasterboard, producing useful material fromwaste materials. The use of recycled bassanite enhanced theperformance of all tested soils. Based on test results, thefollowing conclusions can be drawn:[1] The compressive and splitting strengths increased asrecycled bassanite content increased. The content ofrecycled bassanite had a more significant effect on thecompressive strength compared to the splitting tensile
strength.
[2] The use of recycled bassanite enhanced the performanceof all tested types of soil. However, high performance
was obtained with only sandy soil.
[3] Unconfinedcompressive
strengthdecreasedsignificantly in the case of water content at the wet sidecompared to other cases of optimal moisture and watercontent at the dry side of the compaction curve.
[4] The effect of curing time for samples treated withrecycled bassanite was significant regarding theimprovement of strength, especially in the early curingdays compared with other curing says. The effect of
curing time was not significant after 14 days of curing.
6. ACKNOWLEDGEMENTS
Funding of this research was provided through a grant fromJapan Society for the Promotion of Science (JSPS) under
project number P-10377.
7. REFERENCES

Figure 9. Ultimate strength against water content for soil
type-B treated with bassanite and tested after 14 days curing.

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http://www.gypsumrecycling.biz/ (accessed on June,
2010)[6] Baghdadi Z.A., Fatani M.N., Sabban N.A. Soil
modification by cement kiln dust. Material Civil Eng. J.
1995, 7(4): 218-222.

[7] Miller G.A. and Azad S. Influence of soil type on
stabilization with cement kiln dust. Construction and
Building Materials J. 2000, 14(2): 89-97.
[8] Sreekrishnavilasam A., Rahardja S., and Santagta M.
2007. Soil treatment using fresh and landfilled cementkiln dust. Const. and Building Materials J. 2007, 21(2):318-327.
[9] Ahmed A., Shehata M. and Easa S. Enhancing theStrength of Roadway Subgrades Using Factory-waste
Roof Shingles. In the Proceedings of 37th CSCE 2008
Annual Conference, 10-13 June, 2008, Québec,Canada, vol. 4, pp. 2546-2555.
[10] Ahmed A., Shehata M. and Easa S. Use offactory-waste shingles and cement kiln dust to enhance
the performance of soil used in road works. Advancesin Civil Engineering Journal, Article ID 143750, 9
pages, doi:10.1155/2009/143750, December 2009.
[11] Waste & Resources Action Programme, (WRAP-1).
Plasterboard technical report: Use of recycledplasterboard in unfired clay-gypsum blocks. Projectcode: PBD6-403, Banbury, Oxon, UK, 2007.
Available: www.wrap.org.uk (accessed on June, 2010)[12] Waste & Resources Action Programme, (WRAP-2).
Plasterboard case study: Recycled gypsum in a novel
cementations material. The Old Academy, 21 HorseFair, Banbury, Oxon, UK, 2007. Available:
www.wrap.org.uk (accessed on June, 2010)[13] Holtz R.D. and Kovac W.D. An Introduction to
Geotechnical Engineering. Prentice Hall, New Jersey,
1981.
[14] Kamei T., Kato T. and Shuku T. Effective use forbassanite as soil improvement materials - recycling ofwaste plasterboard. Electronic Japanese Geotechnical
Journal, 2007, 2(3): 245-252. (In Japanese).
[15] Ahmed A., Ugai K. and Kamei T. Investigation ofrecycled gypsum in conjunction with waste plastic
trays for ground improvement. J. Construction and
Building Materials, 2011, 25(1): 208-217.

687

THE EFFECTS OF ENVIRONMENTAL ON TODAY CONSTRUCTION
INDUSTRY
Hooman Abadi, Rouzbeh Abadi, Poona Abadi
Universiti Kuala Lumpur (Unikl) /Business school, Kuala Lumpur, Malaysia,
Universiti Teknologi Malaysia (UTM), Faculty of Civil Engineering, Kuala Lumpur, Malaysia,
Universiti Teknologi Malaysia (UTM), Faculty of Management, Kuala Lumpur, Malaysia
hooman_abadi@yahoo.com
rouzbeh.abadi@me.com
abadi.poona@gmail.com
ABSTRACT:We know three degrees for architecture usefulness, stability and resistance..
Of collection what address to define a correct architecture design from the past to now and can be concluded thatbasically a good design has three characters of simplicity, unity and necessity based on what considered as final
aims in construction design. On the other hand, the basic needs that a construction system must answer it includes
sat and economy function. The construction success rate to meet these needs have been investigated respecting
defined characters. So a plan that supplies the application needs practically and economically, just has answered atleast ``to expectations that there is of a plan.
This article indicates that although the aim to design structure is to generate structure forms that as providing
application needs, tolerate the loads and the structure with economic methods and the most high structure outcomes,
but the aim to design the structure in addition to achieve to finally it is addressed the role of structure engineering art
in forming the new form in the architecture of today's world and the necessity at evolving at the architecture
instruction in the construction industry in order to achieve this goal, as is explained analytically.
Keywords: Construction Industry, Architecture, Challenges, Requirement Effects, Process Design
INTROUDUCTION
Generally, in people’s mind, structural engineering islimited to understanding mathematical concepts,
structure analysis, descriptive visualization of structural
systems and designing them. Most of the people thinkthat task of a structural engineer is to answer needsof an architectural design for creating a structural
system that while can withstand entering forces havepredictable behaviors in different conditions of loading
and also have necessary safety, a process of design thatneeds vast amount of calculations for determining
tensions and deformations. Although structural
engineers really doing these calculations, but theyshould have a very much more portion than only
calculation of structures in forming a modern
architectural plan and this should be noticed as a partof their training while study civil engineering. The aimof structural design is to create strong buildings that
fulfill considered functional requirements as well as
aesthetics needs. Designer should not only consider avast domain of various choices but also should as wellaccurately study behavior of these systems in various

loading conditions and should be sure of securing otherrequirements of the plan.
In this article, role and fundamental effect of structural
engineers in forming a correct, good, and acceptablearchitecture will be analyzed and possible solutions
for realizing the suitable real role of a structural
engineer in designing form and forming various
elements of a building will present. Furthermore,
the effect of innovativeness of structural engineers andutilizing structural engineering knowledge in presenting
structural forms that provide architectural requirements
and aesthetic factor needs of a building will be
discussed in details and needs of noticing this matter inarchitectural training of civil engineering in accordance
to realizing this goal will be stated.
Objectives of structural Design:
Various definitions have presented for a proper, correctand excellent design. Usefulness, stability and beauty,
from whatever has been stated in defining a excellent
design from past, until now, and based on what is
considered in structure design as a final goal, it can beconcluded that basically a good design has below
characteristics:

688

a- Simplicity
b- Unityc- Necessity
Each of these characteristics should be considered in all
stages of design and even in stages of determining
details and designing building elements should beobserved accurately.
Simplicity:
This principle is vastly considered in all variety of arts
and is a common principle in various filed of designingfrom architectural design to industrial design, and its
result is relation and comfort feel that resulted from theintensive effort of designer to realize this very importantprinciple. Structures that in them securing needs andrequirements and comfort of users realize to very hardly
don’t have simplicity. Being simple does not mean
being primitive; structures that match requirementsof the plan and are simply practicable and also does nothave calculation and analysis complexity can be defined
as a structure having a simplicity characteristics.
Unity:
Unity of a plan means coordination, integrity, and being
coherent, not being a unite object. Definition of unity ina flyover bridge is very unusual than the definition of it
in a power plant. These buildings consist of unusual
elements, which have to place together to realizea resolved goal. If elements of a plan have designed forachieving a decided goal, unity will be visible in allelements of the plan. A type of artificial unity is therethat should be avoided. A set of improper, incoherent,
and imbalanced elements can possibly put together but
when this set cannot achieve a unique goal, the result
cannot be something other than unacceptability ofthe plan.
Necessity:
This principle is somehow the key solutionof designing, bet result can be achieved when all
the predicted requirements of a project realized.
This principle derived from noticing innate needsof human. Human mind feel comfort while accepting
something, which is needed, and in presence of extra
unnecessary dummies feels unrest.
These principles (simplicity, unity and necessity) should
not be considered only in design stage of a plan; thesecharacteristics are as factors that a good plan can beevaluated according to them. Teaching these threecharacteristics of simplicity, unity and necessity in civilengineering subjects is one of the most important
requirements to realize such these goals in plans andprojects and this matter needs a new view in educational
planning of this field and creates a change in noticing
architect with modern definition in a civil engineering
field.
Performance, Safety, economy:

Generally, basic needs that a structural system has toanswer them are performance, safety, and economy.
The success rate of building in realizing these needs canbe evaluated based on the stated characteristics.
Therefore, a plan that secures functional needs
practically and economically has just answered
minimum expectations from a plan. Heavy dutyof designing for achieving simplicity, unity, andnecessity make it mandatory to put maximum efforts forachieving best solutions and realizing thesecharacteristics.
Structural Engineering, conception and expectations:
Structural engineering is a modern and rather
a new phenomenon. For a long time, there was nodifference between designers, architectures, andengineers, and even between engineers and builders.
Differentiation of various specialties and separating
duties of structural design from architectural design anddefining structural engineering and determining limited
and special duties for structural engineers in terms ofstructural calculations created a non-innovative imageof structural engineers to the extent that structural
engineers considered as people who just likemathematical science and complicated calculations
related to forces and deformations; and because of thistoday structural engineer have lots of problems
to prove their imagination and innovation ability.
For most of the people and even specialists in general isvery difficult to recognize a real role of these engineers.
Main part of engineering work will be hidden andbecause of that understanding, it is not simple.
In this situation, it is not strange if architectures see
structural engineers of the plan as a technician.
Exchanging ideas is a necessity for a successful
cooperation between an architect and a building
engineer. Engineers have to analyze behavior andreaction of a building in various conditions of loading.
This responsibility make building engineers moreconservative compared to those who don’t haverequired technical knowledge and do not understandthe sensitivity role of engineers and difficulties. Abovestudies shows that the necessity of change in theacademic plan of civil engineering field and also
noticing this matter that building engineers should befamiliar with successful classic and modern patternsof art in the design field just like architectures, artists,
and musicians. Building engineers can achieve very
high imagination and understanding power if they cangain this ability to describe why a special structure
designed perfectly and why other structures does nothave a good design.
Structural Engineering Skills:
Generally, for someone who is not a structure engineer
is difficult to understand necessary skills inthe structural design process. This matter is because689

the only remaining footprint of a designing activity thatpeople will see is completed building that normally astructure of building is hidden inside it. Recognizing
structural engineering skills includes what structural
engineers know and activities that they do during thestructure design process. These skills can be effective inachieving path of aesthetic principles of structure.
Below points include a list of these skills that someparts of it is already predicted in the academic plan ofcivil engineering, and some parts have to be noticed in arevised syllabus of this filed:- Predicting possible loading of a building and roleof reactions of them
- Studying possible ruptures of a structure under effectof forces from both behavior of consisted elements andgeneral stability of the whole structure against forces
as a unit set- Ability of evaluating forces to a structure from
qualitative and quantitative point of views- Having
comprehensive
knowledgeaboutcharacteristics and behaviors of materials andstructures- Ability of matching thinking and view point method tomaterials, structure, and its behavior according todifferent situations- Ability of creating a mathematical model of structureusing geometry, algebra, trigonometry and probability,etc.
- Ability of creating a process for designing and building
structures that are not implemented yet.
- Building sample models in smaller dimensions forevaluation and testing structural theories- Interpretation ability of achieved results of built
samples for utilizing in real dimensions- Choosing practical factors based on evaluations ofa structure and determining limits that value andquantities should not cross them.
- Understanding mutual effect of requirements andcriterions of the plan (such as cost of building,infrastructures of building, services, useful life span,
durability, resistance against fire, and, etc.) on building
structure and vice versa- Evaluating ability and possibility of building proposed
plans- Proposing plans that can be implemented simply andwith lower cost.
- Understanding this matter that in which time and withwhich situation structure design has completed
properly.
Architects and Engineers:
Although structure of building and its function seems
separated from each other but constantly a structure hasa determining role in architecture of building. Structure,
firstly, is compulsory part of a building that cannot bedeleted and secondly, structure should constantly follow
nature and physics rules and therefore, cannot secure
every single demand in architectural design and thirdly,
although structure is a essential part of each building but
generally is hidden and in most of the cases, it seems so
it does not have effect on architecture of building andfinally, building structure have some cost. Howsoever,cost of structure building compared to other costs ofbuilding is not major cost, but even so, it cost money
and generally in building's cost of creating a structure is
one-third to be one-fourth of whole cost of the building.
In some buildings such as bridges or halls with higher
numbers of gates, structure cost is the most part of costof building. Structure is a factor that can create
disagreement between architect and building engineer.
Today a good architect should have general knowledge
about behaviors of structures. Moreover, he/she should
be an artist, bound and committed to observing
principles of architectural design and aesthetical factors.
A plan is successful if its architecture understands its
structure behavior properly, and its structural engineer
has proper knowledge and be committed to architecturaldesign and aesthetical factors. Finally, architect is theleader of a team that is responsible for design andimplementing of a building and all responsibilities andalso credit of success of the project in this section will
be granted to him/her.
A result derived from being knowledgeable andcommitment of a structural engineers on building
design, and aesthetical factors is clearly evident
the need of change in the academic plan of architecture
training in a civil engineering field to achieve abovementioned goals as studying notable architectural andengineering works of the world obvious necessityof such this change.
CONCLUSION
As it has shown in “Objectives of structural design”
section, before anything, building design should berecognized as a joint activity and a group work withparticipation of architect, and structural engineer andtherefore, relating engineering plans to specific persons
is difficult. A modern and innovative design fora building is possible when a process of design andformation of different stages of the plan is resulted of
consultation, cooperation, and effort of an architect,building engineer, infrastructure engineer and even theirbuilders in this path. This matter manifest necessityof teaching required fundamentals for being sure of
realizing such these goals in projects and plan to include
three characteristics of simplicity, unity, and necessity
and also dominant principles on combined realizationof three principles of performance, safety, and economy
in a civil engineering field.

690

Mentioned subjects in “Structural Engineering,
conception and expectations” section shows thatstructural engineer is a member of design and building
team that has extensive knowledge about behaviors ofstructures under effect of different loads. And also
understand production and build process properly.
Structure engineer in the process of formation of anarchitecturalplanshouldtakemonumental
responsibility of structure design, and implementing itin form of building calculations, designing structure
form, designing elements of structure and joints of itwith realizing designing criterions, necessities andlimitations of the plan in various dimensions. From
studying subjects of this section it can be concluded thatprincipal goal of architectural design should be creatingsuch forms that secure operational requirements ofthe building and also be successful in terms of aesthetic.
The main goal of designing structure should be creating
structural forms that while securing functional
requirements can withstand applied loads in best way
and economically. In close cooperation that should bepresent between architects and structural engineers,
the duty of structure engineers is to create plans thanfrom functional view point answer requirements of theplan and from technical view, point should be
implementable, economic and having necessary
conditions and moreover, from the desirable exteriorof building view point complete aesthetical feelingof an architect. Difference between different and yetvery important roles of architect and engineer should
understand well to recognize the real value of work ofeach one of them. Finally, relation between design andconstruction can only change to a positive andconstructive competition that this recognition secured
properly. Study in this section has shown a necessity of
the academic plan of civil engineering and noticing thismatter that civil engineer should be familiar with thesuccessful patterns in the field of classic and modern
design art like architectures and artists. Structural
engineers should gain high level of imagination andunderstanding exact behavior of structures and not only
their precise calculations, and then they will have thisability to understand suitably how a specific building
can answer requirement and criterions of architecturaldesign with proper and correct design while observing
principles of stability of structure. “Structural
engineering skills” and “role of architects andengineers” sections show that the art of structural
engineering is using calculation and design theories
of structures for transforming ideas and concepts into
buildings that can function correctly, utilize materials
economically and answer positively to aesthetical needsof a building. In this situation, structural engineer will
perform its job to reinforce architectural design of thebuilding. Studies in this section have shown that

“Structural engineering is an art” which its goal is tocreate buildings that can answer functional
requirements,
aestheticalfactors,
requirementsof function period, and limitation of conditions related
to construction of building. Structural engineer in thispath not only should consider principles of structuraldesign in choosing the most proper structural system but
should accurately study and evaluate the functionality ofthese systems in various situations for achieving
requirements and mentioned criterions so the processof architectural design and structural calculations reacha unique and common result that can secure all
the requirements of the plan. Study a list of skillsof the structural engineers shows that some parts of it isalready present in the academic syllabus of civilengineering subjects and some parts too should benoticed in a revision of the contents of this field. Thismatter needs a modern view point to academic plan ofcivil engineering filed and create a change in noticing
architecture with a fresh definition in this filed.
REFERENCES:

691[1] Salvadori M. and Levy M., 1981, “Structural
Design in Architecture”, Cliffs, N. J., USA,
Prentice-Hall Inc.
[2] Frampton K., 1985, “Modern Architecture (ACritical Histoty)”, Thames and Hudson
Limited.
[3] Mainstone R.J., 2002, “Developments in
Structural Form”, Cambridge, the MIT Press.
[4] Margolius I., 2002, “Architects + Engineers =Structures”, Willey - academy, London.
[5] Davies C., 1991, “High Tech Architecture”,
London, Thames and Hudson Ltd.
[6] Zunz J., 1992, “Mirror on the Wall, How Fair
Is the Engineering Image?” London, Arup
Journal.
[7] Snelson K., 1989, “The Nature of Structure”,
New York, The New York Academy of
Science.
[8] Siegel C., 1975, “Structure and Formin
Modern Architecture”, Huntington,
New York, Robert E. Kreiger.
[9] Salvadori M., 1990,“Why Buildings Stand Up,
The Strength of Architecture”, New York,
W.W.Norton & Company Inc.
[10] Rice P., 1992, “Royal Institute of British
Architects (RIBA) Gold Medal Speech”,
London.
[11] John Chilton, 2000, “Space Grid Structures”,
Oxford, Architectural Press.
[12] Roth L.M., “Understanding Architecture”,
Herbert Press, 1993.

[13] Golabchi M., 1994, “Selection of Appropriate
Structural Systems for Large Spans”,
Department of Civil Engineering, University ofLeeds, England.
[14] Golabchi M., 1995, “The Art and Practice ofStructural Engineering”, 4th Civil Engineering
Seminar, University of Manchester, England.

692[15] Berger H., 1996, “Light Structures – Structuresof Light”, Basel, Switzerland, Berkhauser.
[16] Cowan H., 1971, “Architectural Structures”,
New York, Elsivier.
[17] Huxtable A. L., 1960, “Pier Luigi Nervi”, New
York, Braziller.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

A Case History of Quay Wall Failure and Remedial Design

C. S. ChenSSP Geotechnics Sdn Bhd, Malaysia
ABSTRACT
A case history of investigation into the failure of a precast reinforced concrete quay wall and the remedial design ispresented in this paper. The quay wall was about 6.5m high. The soil investigation carried out during design stage
indicated that the subsoil below seabed mainly composed of soft clay layer. To enhance the stability, all soft clay was
designed to be removed and replaced by granular material. The quay wall was designed to be founded on top of thisgranular material slope which to be protected by amour rock revetment. However, a stretch of the quay wall moved
excessively during the backfilling works. Investigation was carried out after the incident. Boreholes were sunk toinvestigate the subsoil condition and the results showed that there was a localized very soft clay layer between the
backfilled granular soil layer and the hard soil layer. Few remedial options were considered and stone columns option
was adopted. The quay wall was successfully re-built after the stone columns were installed.
Keywords: Quay wall, stone column, soft clay, slope.

1. INTRODUCTION
A harbour basin had been planned for a marina project in amixed development on a piece of coastal land. The existing
ground was generally on low lying area with most of the time
submersed under water. In order to allow boats and yachts tosail from the open sea into the marina project site and berthinside, about 8 to 9m deep dredging from the existing groundsurface was required to form a basin. For the surrounding areaof the basin, reclamation was required to form a platform
higher than the tide level. Quay wall of about 6.5m high was
designed to be constructed at the perimeter of the basin toretain the backfilled soil. Fig. 1 shows the proposed layout
plan and Fig. 2 is the model of the proposed marina project.

mainly composed of boreholes was carried out. The subsoil
information gathered showed that there is a thick layer of softcohesive soil below the existing ground surface. It was
decided to remove the soft soil and replaced by granularmaterial which mainly consisted of sand in order to ensure thelong term stability of the basin slope and sufficient bearing
capacity for the proposed quay wall.

CUT AREA

Fig. 2 Model of the proposed marina project.

PROPOSEDQUAY WALLPROPOSEDSHORE PROTECTION

HARBOUR BASIN(DREDGED TO -8.5m LSD)PROPOSEDSILT CURTAIN

SE

A

However, a stretch of the quay wall moved excessively
during the backfilling of soil behind the quay wall.
Investigation was carried out to identify the possible causes offailure and to gather additional information for remedial
design. This paper presents the subsoil information at sitegathered from the soil investigation, findings and possible
causes of failure from the failure investigation and theremedial design.

Fig.1 Layout plan of the marina project.
During the planning and design stage, soil investigation

693

For ease of construction, quay wall will be precast at factoryand delivered to site during construction. Fig. 4 shows the
typical design cross section of the proposed quay wall. As itwas expected that the quay wall could be founded on the soft
clay layer, stability of the quay wall was a concern in thedesign. Stability analysis carried out indicated that with theremoval of the soft material and replaced with granular soil,
the wall will be stable. It was decided to remove all soft
material beneath the quay wall foundation. To prevent
scouring or erosion takes place, the granular slope wasprotected by a layer of rock revetment.

2. SUBSOIL CONDITION
Soil investigation mainly consisted of boreholes was carriedout during the planning and design stage to gather the subsoil
profile and engineering properties. Typical subsoil profile atthe proposed basin area is shown in Fig. 3. In general, soft
clay was found below the existing seabed with thickness
varied from few meters to about 10m. Liquid Limit andPlasticity Index for this soft clay layer are in the range of 40%
to 101% and 16% to 77% respectively. A layer of loose sandof silt could be found below the soft clay layer. The very
dense sandy soil layer or hard bearing stratum is generally atabout 13 to 15m below the existing seabed.
0

4. FAILURE OF THE QUAY WALL

Fig. 3 Typical subsoil profile.

Dredging work was carried out immediately after the design
was completed. The basin was dredged to the designed level.
For areas below the quay wall and the slope, over-dredged
was required in order to remove all soft material which may
affect the stability of the quay wall. Once the removal of soft
soil was completed and granular material was filled to thedesigned founding level, quay wall was installed andbackfilling started subsequently.
However, a stretch of the quay wall moved excessively into
the basin before the backfill reached to the designed platform
level. The backfill material behind the quay wall wasimmediately removed to mitigate the wall movement.
Investigation of the failure was carried out after the incident.
Fig. 5 shows the failure condition of the quay wall.

3. DESIGN OF THE QUAY WALL

5. CAUSE OF FAILURE-1Existing Ground Level at -0.8 mLSD

-2

Very soft silty clay
SPT-N = 0
LL= 40~101%
PI = 16~ 77%

-3
-4
-5

Ground Level (m LSD)

-6
-7
-8
-9
-10

Loose sandy silt and silty sand
SPT-N = 6 ~ 9
LL = NP~ 33%
PI = NP~ 10% (*NP=non-plastic)

-11
-12
-13
-14

Very dense silty or gravelly sand
SPT-N >50

-15
-16
-17
-18
-19
-20
0

10 20 30 40 50
SPT-N (blows/300mm)

The design of the quay wall was reviewed. It was found thatthe design should be adequate. The construction method andsequence of work were also being reviewed and it was
concluded that these should not be the cause of failure.
Initially it was suspected that the rock revetment could havebeen washed away by the current which undermined thematerial below the quay wall and caused the wall to moveexcessively. However, probing results indicated that rockrevetment and the slope profile still maintained. The
abnormal observation was that there were signs of heaving at

The existing ground was a low lying mangrove swamp landwith the ground levels were mostly in the range of +1m LSD(Land Survey Datum) to -1m LSD. At normal and high tideperiod, the ground was submersed underwater. The design ofthe basin required to excavate the existing ground to thedesigned level of -8.5m LSD. The land surrounding the basin
was required to be backfilled to a higher platform level ofabout +3.1m LSD. Quay wall of about 6.5m height was
designed to retain the backfill soil.
PROPOSED QUAY WALL

OGL

?

?

?

?

?

?

?

?

Fig. 4 Typical design of quay wall
694

?

?

?

the toe of the slope. To gather more subsoil information andengineering properties at the failure zone, boreholes wereplanned and carried out. Fig. 6 presents the gathered subsoil
information. Generally the subsoil profile was consistent withthe design assumption. Backfilled sandy soil was foundbehind and below the founding level of the quay wall.
However, a layer of soft clay was detected in within the
backfilled sandy soil and the medium stiff to hard soil layer.
Stability analysis was carried out based on the obtained soil
profile and the result showed that the slope was unstable. It
was concluded that the failure of the quay wall was mainly
caused by the existence of this soft clay layer. In the design,
this soft clay layer was required to be removed. However, thisrequirement was not fully implemented during the dredging
stage. Although some exploratory holes had been sunk prior
to the dredging work to establish the expected depth for soft
clay removal, Part of the reason was that the workers
followed the estimated depth of removal which wasestablished from the results of exploratory holes carried out
prior to the commencement of dredging work.

Fig. 5 Conditions of failed quay wall.

Also, it was difficult for a non-experienced engineer todetermine whether the subsoil is soft or medium stiff at site todecide the actual depth of removal. This incident could havebeen avoided if (a) there was an experienced site staff toinspect the dredged subsoil and determined the actual
required depth for removal of soft soil and (b) carry out someverification measures to ensure that all soft soil had been
removed prior to the backfilling work.
6. THE REMEDIAL DESIGNThe failure occurred at localized stretch only. For quay wall atother area, the backfilling work had reached to the designed
platform level and the wall was found in stable condition. In
selection of suitable remedial measure, one of factors needed
to be considered was not to disturbed the already constructed
quay wall.
Few options had been considered. The advantages anddisadvantages of each option are discussed as follow:
6.1 Replacement Option
This option was similar to the original design, i.e. to re-dredge
the soft clay which was found from the investigation andreplaced with granular material. The advantage of this option
is that the staff at site was fully aware of the failure cause andthey will ensure all soft soil underneath to be removed. Also,
they were familiar with the sequence of work. The
disadvantage was this will affect the stable quay walls at bothsides of the failure stretch.
6.2 Ground Improvement Option
This option is to improve the strength of the soft clay layer inorder to enhance the global stability of the quay wall. A wide
range of soil improvement methods [1] is available. For thissite, vibro stone columns, cement mixed columns andgrouting are more suitable. Surcharging method with orwithout Prefabricated Vertical Drains was not considered as
this option required longer time. The main advantage of this

PROPOSED QUAY WALL

OGL

Fig. 6 Subsoil profile from the investigation results
695option is minimum disturbance to the adjacent quay wall. The
disadvantage is that the cost is usually higher as comparedwith the first option. Vibro stone column was the better option
as compared with cement mixed column and grouting after
consideration of the availability of materials, experiences andmachinery.
6.3 Structural Option
This option was to use structure such as steel sheet piles orcontiguous bored piles to stabilize the slope. The advantage isthe construction is fast and straightforward. However, thisoption is the most expensive option.
6.4 The Vibro Stone Column Remedial Works
In order to minimize the disturbance to the adjacent quay walland to maintain the cost within acceptable level, vibro stone
column option was selected. The design of stone column canrefer to [2]. It was decided to adopt 800mm diameter stone
columns with center to center spacing of 1.8m in triangulargrid after analysis was performed to improve the soft clay soil
underneath. Fig. 7 presents the typical section of the stone
column option.

Fig. 8 Stone column installation work at site

PROPOSED QUAY WALL

OGL

Fig. 7 Typical design of stone column option

Fig. 9 Condition of the successfully built quay wall

The remedial work started by removing the affected quay
wall. A working platform was then prepared to allow the
movement of the machinery and installation of stone columns.
Fig. 8 shows the stone column installation work carried out atsite. The vibroflot with water jetting created an open hole tothe designed depth. A long arm excavator was used to feed in
the required stones and the vibroflot started to compact thestones until the stone column was formed. The quay wall was
successfully built after the completion of the installation ofstone columns. Fig. 9 shows the quay wall condition a year
after the completion of the construction work.

8 REFERENCES[1] Mitchell J.M. and Jardine F.M, A guide to groundtreatment. CIRIA publication C573, 2002.
[2] Priebe, H.J, “The design of vibro-replacement”. GroundEngineering, Vol. 28, No. 10. 1995, pp.31-37.

7 CONCLUSION
A stretch of quay wall moved excessively before the backfill
behind the wall reached to the design level. Investigation
results show that there was a layer of soft clay below the quay
wall which was required to be removed in the design. Due tothe low strength of this soft clay layer, the quay wall was
unstable. Stone column option was adopted to improve thesubsoil strength. The quay wall was successfully built after
the stone columns were installed.
696

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Environmental Economical Efficiency in Treatment/Reutilization ofConstruction Sludges considering Delay in Process Flow
Shinya Inazumi1, Hiroyasu Ohtsu1 and Takayuki Isoda1
1
Kyoto University, Japan
ABSTRACTIn Japan, although the recycling of waste has been promoted, there may be various obstacles. In this study, we focus on
construction waste, more specifically, construction sludge, which is less advanced in terms of recycling. Processing
cost for materials recycled from construction sludge is higher compared to the cost of mining virgin materials. There aresome additional difficulties including delays in schedules due to the time required for processing and the balance
between supply and demand. Therefore, we quantitatively assess the benefit of recycling by adding the time factor to
the evaluation model with consideration for cost and environmental load. As a result, it is shown that the bottlenecks ofconstruction sludge recycling are higher cost than the cost of mining of virgin materials, high heavy metal contents inthe construction sludge, and delays in the recycling process.
Keywords: Environmental Economics, Environmental Impact Assessment, Construction Sludge, Construction waste,
Processing, Recycling
1. INTRODUCTION
Although recycling of construction waste is advancing, there
are still many problems to be solved. One of the main factors
limiting recycling of construction waste is the higher unit
price of recycled materials compared to that of virginmaterials because of the necessity of processing for recycling.
In addition, it takes time for such processing for recycling,
and if recycled materials are used, the construction schedules
could be extended, resulting in a possible increase in theconstruction cost. However, we can look upon these problems
as being the result of considering recycling only in terms ofcost. The purpose of recycling consists of enabling society toundertake sustainable development through environmental
preservation. Thus, in evaluating the business potential ofwaste recycling, not only the required actual cost but also theeffects on the environment should be considered. In otherwords, it is important that business activities related to waste
recycling should be evaluated according to environmental
economics, which employs a well-balanced consideration
between actual cost and environmental effects by accounting
internally for environmental impact over the life cycle basedon the concept of environmental economics.
In the previous studies regarding construction waste, we
examined the environmental impact assessment andenvironmental accounting approach to evaluate recycling of
construction waste by environmental economics in which
environmental load converted to environment cost in addition
to direct cost is considered [1]-[2]. In this study, we focus on
the extension of construction schedules caused by process
delay brought about from processing for recycling that is oneof the obstacles for recycling of construction waste.

Specifically, regarding recycling of construction waste, we
use environmental impact assessment and the environmental
accounting approach to perform environmental economic
evaluation with consideration of process delays in processing
for recycling. In this study, among construction waste that is
supposed to be a recycling resource, we look at construction
sludge, which is less advanced in terms of recycling.
2.

DEFINITION OF CONSTRUCTION SLUDGE

2.1 Present Situation of Construction Sludge
According to the Ministry of the Environment, the data forindustrial waste by industry (2008) shows that theconstruction industry discharged 700 x 104 tons, 18.9% of thetotal, which was the third largest amount [3]. Similarly,
according to the field survey about byproducts ofconstruction by the Ministry of Land, Infrastructure,
Transport and Tourism, the data for construction waste byitem (2008) shows construction sludge was 450 x 104 tons [4].
Construction sludge did not constitute such a large proportionof the total construction waste, but the data for the final
disposal by item of construction waste (2008) shows that
construction sludge was 67 x 104 tons, approximately 15% of
the total construction waste, 402 x 104 tons [4]. When it iscompared with the previous field survey of construction
byproducts (2005), it can be considered that the situation
regarding the final disposal of construction sludge improved,
because the final disposal of construction sludge was 192 x
104 tons and it constituted approximately 30% of the total,
but it was still a large proportion [5]. On the other hand, as forthe recycling of construction waste in 2008, the total
recycling rate was 93.7%, and the recycling rate ofconstruction sludge was 85.1% [4]. This was also an
improvement from the previous survey. However, therecycling rate for construction sludge was still smaller than697

the rate for the other construction waste. The recycling rate isdefined as a percentage of the recycled amount or the reduced
amount against the total amount. And the effective
availability rate of displaced soil at construction is defined asa percentage of the displaced soil at construction against the
total soil used for construction.
2.2 Obstacle for Recycling of Construction Sludge
It is preferable to use the discharged construction sludge
during the construction period. However, according to thefield survey of construction byproducts in 2008, thepercentage of the use during the construction period was only
2% [4]. This was because there are few facilities available forrecycling and it required long distance transportation between
a construction site and a processing facility in many cases. In
other words, transportation cost increases and furthermore,lots of air pollution is produced. In addition, for recycling of
construction sludge there should be construction sludge in thefirst place. Even if construction needs recycled materials,
recycled materials cannot be supplied without construction
sludge at the processing facility. In addition to thesedisadvantages, recycling takes time and it may cause a delay
in construction. If the construction schedule is delayed, it willlead to an increase in costs, such as labor cost or lease
payments for the equipment. Thus, constructors are forced touse virgin materials instead of recycled materials. The
possible delay in the construction schedule is one of the
obstacles for the recycling of construction sludge.
3.

ENVIRONMENT IMPACT ASSESSMENT AND
ENVIRONMENTAL ACCOUNTING APPROACH

3.1 Quantification of Direct Cost
For recycling of construction waste, an inevitable cost
including transportation cost and processing cost forrecycling is required. In this study, we refer to these types ofcost as direct cost. When we roughly classify them, we havean initial cost and running cost (including processing cost,transportation cost and storing cost) [6]. The direct cost canbe calculated by the Eq. (1). We apply the Eq. (1) to the
manufacturing of virgin materials, manufacturing of recycledmaterials and disposal. The elements in Eq. (1) indicate unit
price of unit construction work.
C = CI + CS + CT + CK

(1)
CS = WS CT = WLT C K = WDK
Where, C: direct cost (yen), C I : initial cost (yen), C S :processing cost (yen), C T : transportation (yen), C K : storing
cost (yen), W: mass (t), S: unit price required for processing ofeach material (yen/m3), L: transportation distance (km), T:
unit price required for transportation (yen/t⋅km), D: storingdays (day), K: unit price required for storing each material(yen/t⋅day)
3.2 Quantification of Environmental Cost
In this context, it is necessary for people to understandenvironmental value through its conversion into monetary

value. In this study, we refer to detected environmental load
converted into a cost-based figure just like direct cost as
environmental cost.
In the evaluation and comparison based on environmental
economic efficiency performed in this study, Eq. (2) showsthe environmental cost element to which we pay special
attention. We apply Eq. (2) to the manufacturing of virginmaterials, manufacturing of recycled materials and disposal.
Each element in Eq. (2) indicates the unit price of unit
construction work. Here, we refer to the results of the study of[7]-[8] and other previous studies [2] for extraction ofenvironmental cost elements.

E = EO + ET + EC1 + EC 2 + EC 3 + EC 4

(2)

Where, E: environmental cost (yen)，E O : environmental
cost accompanied with operation of facilities (yen) ， E T :environmental cost accompanied with transportation (yen)，
E C1 : environmental cost related to public functions such as
forests (yen)，E C2 : environmental cost related to impact on
ecological system (yen)，E C3 : environmental cost related to
exploitation of natural resources (yen)，E C4 : environmental
cost related to worsening of residential environment (yen).
3.3 Business Effect of Recycling of Construction Waste
It is thought that recycling of construction waste contributes
to the slowdown in the decrease of the remaining capacity of
the final disposal sites through reduction of waste, and to cost
reduction of waste disposal [9]. In this study, we regard thesebenefits of recycling as business effects caused by reductionof cost required for waste disposal. The evaluation of thebusiness effect is calculated by multiplying the amount ofconstruction waste by the unit price of final disposal as shownin Eq. (3). However, since the business effect means cost
reduction, the calculated result should be a negative value.

B = −W × S(3)

Where, B: business effect of recycling of constructionwaste (yen), W: amount of construction waste (m3), S: unit
price of final disposal (yen/m3).
3.4 Total Cost
The environmental assessment and environmental accounting
approach defines the total cost shown by Eq. (4) as one of theindexes of environmental economic efficiency. The total cost
is considered as a social cost with consideration of theenvironmental impact.

T =C+E+B(4)

Where, T: total cost (yen) ， C: direct cost (yen) ， E:
environmental cost (yen)，B: business effect by recycling(yen).

698

3.5 Consideration of Delay in Processing Process
One of the obstacles for recycling is that the supply ofrecycled materials does not fit well with the construction
schedule [5]. This is because the recycling of constructionwaste has some uncertainties including that constructionwaste is not constantly supplied to the disposal site, that it
takes time for recycling to be certified because the method ofrecycling is complicated or imprecise, and that the timerequired for processing fluctuates because the quality of
construction waste varies depending on the site. As acountermeasure to any delay that may occur in the process of
disposal, storage of waste at the stock yard has been proposed.
In this study, we assume soil material used for soil structures
can be stored until the specified storage capacity of the site isreached. The number of days for storage is determined by theamount that can be dug in a day in case of virgin material orby the amount that can be manufactured in a day in case ofrecycled material. If the material stays at the stock yard long,the cost for storage increases. The number of days for storage
is defined by Eq. (5).

D = W / Wi

(5)

Where, D: the number of days for storage (day), W:
required amount of soil for a soil structure (m3), W i : amountof soil (virgin or recycled) that can be produced in a day(m3/day).
3.6 Evaluation by Monte-Carlo Simulation
There are various approaches to analyze the recycling of
construction waste while considering every uncertainty. In
this study, we adopted the Monte Carlo simulation. The
Monte Carlo simulation is a technique to comprehend all thepossible combinations by changing every uncertain element
at the same time. By using the Monte Carlo simulation, we
can obtain frequency distributions of the output and apply the
results to evaluate business risk.
4.

EVALUATION OF ENVIRONMENTAL
ECONOMIC EFFICIENCY OF CONSTRUCTIONSLUDGE RECYCLING

In this chapter, we apply the environmental impact
assessment and environmental accounting approach torecycling of construction waste and perform evaluation andcomparison of virgin materials (not regenerated soilmaterials) and recycled materials (regenerated construction
sledge) based on environmental economic efficiency.
4.1 Assumptions
The following are assumptions for evaluation and comparison
for recycling of construction sledge based on environmental
economic efficiency.
(1) The amount of soil for the target soil structure is
3,000m3.
(2) For each element of uncertainty, the minimum, median(base case) and maximum are specified as shown in

Table 1. We decided the medians either from theaverages of the values in literature searchor based onwhat we obtained from interviews [7], [10]-[11].
(3) The processing for recycling of construction sledge is astable treatment.
(4) As shown in Fig. 1, when the processing cost increases,
the amount of materials and energy required forprocessing as well as the environmental costaccompanied with operation of the facility (processing)increase.
(5) For digging, the rate of change is assumed. When soil is
dug up, gaps appear in the soil. As a result, the mass ofthe soil on the ground or after digging becomes larger
than the mass after compaction. This rate of change insoil mass is defined as the soil conversion factor.
(6) The quality of virgin material and recycled material is
assumed to be same and it is assumed that there is nodifference in construction technique.
(7) The material (virgin and recycled) for a soil structure isused 100% without being discarded.
(8) The possible heavy metal contained in construction
sledge is lead (Pb).
(9) When the cost of processing for recycling increases, theheavy metal content in the recycled soil will be reduced
as shown in the Fig. 2. In this study, we assume theremoval of heavy metals is included in processing forrecycling.
(10) Virgin material and recycled material are stored in thestock yard and the construction work starts when theamount in storage reaches the amount required for theconstruction work.
(11) Recycled material receives caring at the stock yard forthree days.
Heavy metals contained in construction sludge can be an
obstacle to recycling, but in fact there are few examples ofheavy metals contained in recycled materials [4]. At the sametime, the Guideline for construction sludge recycling [10]under the editorship of the Ministry of Construction does not
specifically mention the measures related to heavy metals
contained in recycled materials. Thus, we use two
assumptions for calculating environmental cost. One isrecycled material contains heavy metals and the other isrecycled material does not contain heavy metals. In the latterassumption, we do not consider environmental cost regarding
the ecosystem.
4.2 In Case of No Delay of Processing for Recycling
Figs. 3 and 4 show the results of the Monte Carlo simulation
focusing on total cost (histograms show frequency
distributions). They show two cases: recycled material
contains heavy metals and recycled material does not containheavy metals. In the case of recycled material does notcontain heavy metals, for 96.8% of all the results, theappraisal value became 0 or larger, which means recycledmaterial is superior to virgin material in terms of total cost. In
summary, the environmental economic efficiency of recycledmaterial is better than that of virgin material in most of the699

Table 1 Minimum, medium, maximum value in each
uncertainty factor
Volume of compacted soils(m3)
Volume of mining (m3/day)
Volume of recycling(m3/day)
Mining cost (yen/m3)
Recycling processing cost(yen/m3)
Storing cost (yen/m3/day)
Unit cost of final disposal(yen/m3)
Unit cost of transportation(yen/t/km)
Rate of change volume of
soils
ΔNPP
Heavy metals contents(mg/kg)Mining area (ha)

Min.
value

Med.
value

Max.
value

-

3,000

-

-

600

-

-

600

-

1,000

3,000

5,000

2,000

5,000

8,000

-

1,500

-

4,000

7,000

10,000

58

68

83

1.26

1.47

1.70

0

2

9

0

23.1

150

-

0.1

-

Fig. 3 The result of MonteCarlo simulation when heavy
metals are not considered in case of No delay

Fig. 4 The result of MonteCarlo simulation when heavy
metals are considered in case of No delay
Table 2 Acquired correlations in case of No delay
Heavy metals are
not considered
0.421

Heavy metals
are considered
0.371

Recycling processing cost

-0.602-0.121

Final disposal cost

0.577

0.545

Unit cost of transportation
Transported distance invirgin materials
Transported distance inrecycling materials
ΔNPP
Heavy metal contents-0.072-0.069

0.12

0.101

-0.163-0.119

0.054
-

0.017
-0.599Mining cost

Fig.1 The relation between processing cost andenvironmental cost

Table 3 The degree of correlations±0.7～±1±0.4～±0.7±02～±0.4±0～±0.2

Fig.2 The relation between processing cost and heavy
metal contents
cases. On the contrary, in case of recycled material contains
heavy metals, for 78.3% of all the results, the appraisal value
became 0 or larger, which means the probability that recycledmaterial is superior in economic efficiency to virgin material
is rather high. However, it has been clarified that thesignificance of the evaluation regarding recycling ofHigh correlation
Medium correlation
Low correlation
Little correlation

construction sludge becomes far larger when evaluated bytotal cost considering the load on the environment than whenevaluated only by direct cost.
Yet there is a case in which recycling is not necessary evenwhen evaluated by total cost, and analysis of such cases
should be undertaken. Nevertheless, it takes many labor hours

700

to analyze a great number of uncertainties, which may bringabout an obstacle to recycling because of the cost increase ora delay in the construction schedule. Therefore, we
considered that we should focus on uncertain elements thathave large influences and seek the correlation between
uncertain elements and appraisal value. Tables 2 and 3 show
the acquired correlations and the degree of the correlations.
Here, positive correlation implies that the appraisal value
increases as the value of uncertain element increases, while
negative correlation implies that the appraisal value decreases
as the value of uncertain elements decreases. Table 2 showsthat for recycled material that does not contain heavy metals
the digging cost has the positive correlation of the medium
degree and the processing cost for recycling has the negative
correlation of the high degree, and that for recycled material
that contains heavy metals, similarly, the digging cost has the
positive correlation of the medium degree and the inclusion ofheavy metals has the negative correlation of the high degree.
In addition, for both cases, the disposal cost, meaning thebusiness effect of recycling of construction sludge, has thecorrelation of the high degree. According to these results,
especially digging cost, processing cost for recycling, content
of heavy metals and final disposal cost are important when weanalyze the environmental economic efficiency of recycledmaterials. These four become important indexes in order topromote recycling of construction sludge.
4.3 In Case of Delay of Processing for Recycling
Similarly in the previous section, we conducted anassessment by using the appraisal value which is thedifferential of the total cost between virgin material andrecycled material as the index. When we compared recycledmaterial that does not contain heavy metals with virginmaterial, the probability of the appraisal value being 0 or
larger, meaning recycled material that does not contain heavy
metals is superior, was 76.2%. In contrast, when wecompared recycled material that contains heavy metals with
virgin material, the probability of recycled material that
contains heavy metals being superior became 40.5% (see Figs.
5 and 6) Here, we conducted an assessment of the casewithout considering a process delay of processing forrecycling, and the resultant probabilities are as follows: theprobabilities of the appraisal value being 0 or larger became
96.8% for the case of recycled material does not containheavy metals and 78.3% for the case of recycled material
contains heavy metals. In summary, the result showed
process delay of processing for recycling caused theusefulness of recycling to be reduced.
In case of considering a process delay of processing forrecycling, the usefulness of recycling decreases and moredetailed pre-examination is required. Table 4 shows thecorrelation between uncertain elements and appraisal values.
For recycled material does not contain heavy metals, themanufacturing amount of recycled material as well as thedisposal cost have positive correlations to the medium degree,
and the processing cost for recycling has a negative
correlation to the medium degree. Also for recycled material

Fig. 5 The result of MonteCarlo simulation when heavy
metals are not considered in case of Delay

Fig. 6 The result of MonteCarlo simulation when heavy
metals are considered in case of Delay
Table 4 Acquired correlations in case of Delay

Volume of recycling
Mining cost
Recycling processing
cost
Final disposal cost
Unit cost oftransportation
Transported distance invirgin materials
Transported distance inrecycling materials
ΔNPP
Heavy metal contents

Heavy metals are
not considered
0.556

Heavy metals are
considered
0.518

0.338

0.298

-0.455-0.094

0.468

0.471

-0.049-0.055

0.029

0.037

-0.168-0.159-0.046--0.011-0.484contains heavy metals, the manufacturing amount of recycledmaterial as well as the disposal cost has positive correlations
to the medium degree. In addition, it was found that thecontent of heavy metals has a negative correlation to themedium degree. According to these findings, the stablemanufacture and supply of recycled material has a very large
influence. In order to promote recycling, it is essential toestablish a system in which the stable supply of recycledmaterial can be performed. In particular, information about
the quality and the amount of soil material required forconstruction of a soil structure as well as about the701manufacturing amount at a facility of processing of recycling
should be closely shared among the related parties. This way,
the supply and demand of soil material can be comprehended
precisely, and more flexible use of virgin and recycledmaterials can be performed, so that the bad effects of aprocess delay of processing for recycling of construction
sludge can be controlled.
5.

CONCLUSIONS

The results we obtained are as follows:(1) Regarding evaluation and comparison of environmental
economic efficiency of recycling of construction waste,
we could include the environmental load that was notconvertible into CO 2 in the cost base in addition to theconventional cost depending only on CO 2 emissions.
This way, we were able to combine environmental loadand direct cost, which have different dimensions, andcreate a new index: total cost.
(2) By performing evaluation and comparison ofenvironmental economic efficiency through the Monte
Carlo simulation, we considered the uncertainty ofassumed elements. And we showed that the relative
significance of evaluation regarding recycling of
construction sludge increased when evaluated by total
cost than when evaluated only by direct cost.
(3) We showed the usefulness of recycling decreased
quantitatively by considering a process delay inprocessing for recycling owing to unstable supply of
construction sludge as well as a variance in the periodrequired for processing for recycling.
(4) We clarified that limiting factors for recycling of
construction sludge include the processing cost forrecycling is higher than the digging cost of virginmaterial, construction sludge contains heavy metals anda process delay in processing for recycling.
6.

[6] Matsuo M and Honjo Y, “New View of Geotechnicaland Environmental Engineering -The Efficient Use of
Displaced Soils at Construction-“, Gihodo Shuppan,
1999.
[7] Omine K and Matsuyuki K, “Environmental economic
model for recycling of construction surplus soil andwaste material”, The Magazine of the Japanese
Geotechnical Society, JGS, Vol.51, No.5, 2003,
pp.10-12.
[8] Omine K, “Life Cycle Assessment”, The Magazine of
the Japanese Geotechnical Society, JGS, Vol.55, No.10,
2007, pp.40-41.
[9] Agriculture, Forestry and Fishers Bio Recycling
Laboratory System Team, “Design and Analysis of
Biomass Utilization System”, National Institute for
Rural Engineering, 2006.
[10] Advanced Construction Technology Center, “The
Guideline in Construction Sludge”, Taisei Publishing,
1999.
[11] Itsubo T and Inaba A, “Life-cycle Impact Assessment
Method based on Endpoint Modeling”, Maruzen
Publishing, 2005.

REFERENCES[1] Kunibe K, Itsubo T and Mizuguchi T, “Environmental
Accounting”, Yuhikaku Publishing, 2007.
[2] Inazumi S, Ohtsu H, Isoda T and Shigematsu Y,“Evaluation of social environmental efficiency on
recycling of construction sludges as ground materials”,
JSCE Journal of Geotechnical Engineering, JSCE,
Vol.68, No.1, 2012, pp.163-174.
[3] Ministry of the Environment, “The Report of the
Emission and Disposition of Industrial Waste in 2008”,
Announce Document, Ministry of the Environment,
2008.
[4] Ministry of Land, Infrastructure, Transport and Tourism,“The Report of the Investigation of Construction Waste
in 2008”, Announce Document, Ministry of Land,
Infrastructure, Transport and Tourism, 2008.
[5] Ministry of Land, Infrastructure, Transport and Tourism,“The Report of the Investigation of Construction Waste
in 2006”, Announce Document, Ministry of Land,
Infrastructure, Transport and Tourism, 2006.
702

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Evaluation and validation of elasic-fragile damage modeling for unsaturated
porous media in θ-Stock
B. Gatmiri1,2 , M. Fathalikhani3Center of Excellence for Engineering and Management of Infrastructures (CE-EMI). Department of civilengineering, University of Tehran, Tehran, Iran
2
ANDRA, Direction Scientifique - Mécanique des Fluides et des Solides, Paris, France
3
Department of Civil Engineering, university of Tehran, Tehran, Iran1

ABSTRACTThis paper presents a framework for the modeling of damage in continuous unsaturated porous geomaterials. The
THHMD model developed by Gatmiri & Arson (2008) has been formulated in net stress, thermal stress and suction
independent state variables. The behaviour law stems from both micromechanical and phenomenological concepts.
Correspondingly, the strain tensor has been split into two independent thermodynamic strain components. A crack
related intrinsic water permeability has been introduced.
A numerical modeling is simulated with θ-Stock finite element program [1] to study damage in heated samples of
unsaturated bentonite. In this paper the results of this simulation will be discussed.
Keywords: Unsaturated soil, Thermohydromechanical behavior, Finite element method, Damage modeling,
Multiphase porous media
In this section we just present the main concepts of damage
model in order to help us to deduce and adapt the deductions
to the damage concepts and basic theories.
Use Supposing for example that the material is fractured ina

1. INTRODUCTION
The purpose of this study is to predict the behavior of the
Excavation Damage Zone surrounding nuclear waste
disposals. The geological barriers, often made of quasi-brittle
material like granite or clay-rock, undergo damage during the
excavation phase. Hydro-mechanical interaction may occur
in the neighborhood of the engineered barrier, which is
generally constitute of unsaturated compacted clay.
In Continuum Damage Mechanics, almost all models
developed for unsaturated media are based on the concept ofBiot’s effective stress, stemming from Bishop’s definition ofstress. Some important aspects of the behavior of unsaturated
soils, like wetting collapse, cannot be represented by thistheoretical frame [2]. Alternatively, the THHMD model
involves independent state variables (net stress, suction andthermal stress), in order to emphasize the role of suction
rigidity [3] Waste is a heat source which can generate
traction, and thus cracks; therefore, the design of deep nuclear
waste repositories requires the modelling of the effects ofthermal loadings in the Excavation Damaged Zone (EDZ).
The theoretical frame is briefly recalled in the first part of thispaper, also the numerical parametric study is performed inorder to determine the main parameters controlling thegeneration of damage in THHMD model.

three principal directions, the damage variable Ω can be

written as a diagonal tensor whose eigenvalues
crack densities:

d i represent3Ω = ∑ d i ni ⊗ni

i =1
(1)
Adopting the definition 1 implies that damage can bequantified by three fictive homogenized fracturescharacterized by a normal vectorni

and a relative volume

d i .The stress state variables chosen to formulate the model

are net stress, σ ij′′ = σ ij − p a δ ij , suction s = p a − p w andthermal stress pT [4].
Here, net stress and suction are thermodynamicallyconjugated with mechanical (εM) and suction-related strains(εSv) respectively. Both of these components are split in areversible and an irreversible part:
rev
irr
dε Sv
dε Sv
irr
δ ij + dε Mδ ij , dσ "ij
ij +
3
3
rev
rev
= De (Ω )ijkl dε M
lk , ds = β s (Ω )dε Sv
rev
dε ij = dε M
ij +2. BASIC CONCEPTS OF THHMD DAMAGE MODEL
2.1 A Formulation in Independent State Variables

ni

(2)

Ω is the damage tensor, and De(Ω) and βs(Ω) are the damaged
mechanical and capillary rigidities respectively.

703

2.2 Behavior laws
The following expression of the free energy has been
postulated, on the basis of the approach introduced by Halmand Dragon [5]:1ε Mji Deijkl (Ω ij )ε Mlk
2
1
1+ ε SV β s (Ω ij )ε SV + ε TV β T (Ω ij )ε TV
2
2gSgδ ij Ω ji ε SV − T δ ij Ω ji ε TV
− g M Ω ji ε Mji −3
3

The details of thermo-hydraulic transfers in intact unsaturated
porous may be found in [3].
Moisture flows have been assumed to be diffusive [3]:Vw = −

Vvap =(3),ε ,ε , Ω )provides the whole stress/strain relations. The damage stress

Yd , conjugated to damage, writes:1
Yd ij = − ε M mn
2

∂Ω qp

(8)

σ(T) is the superficial energy of pore water (in J.m-2). DTvap
and DPvap are the thermal and capillary vapour
conductivities, respectively. The water permeability tensor is
split in an intrinsic part and in relative components:

Kw = kT (T ) k r ( S w ) K int (n, Ω ij )(9)

The thermal and capillary relative permabilities kT (T ) andk r ( S w ) are related to heat and to the behaviour of pore fluidsk T (T ) =S w − S w, r 3µ w (T ), k r (S w ) = ()
1 − S w, r
µ w (Tref )In whichµ w (T ) is the dynamic viscosity of liquid water,

S w, ris the residual water saturation degree. The water

(4)

and

The damage yield function is assumed to depend on theggtensile strains involved in g M ε M ij + s ε Sv δ ij + T ε Tv δ ij
3
3
only. The corresponding thermodynamic variable is noted

S

1

−1+
S w = [(1 − S w, r )(1 + (αs ) n ) n + S w, r ]
× exp(d s (T − T0 ))ifS = 1if w

domain:(5)

C 0 is initial damage-stress rate that is necessary to trigger
damage. C1 controls the damage increase rate. The damage
evaluation law is computed by an associative flow rule.

(10)

saturation degree w evolves on a thermo-hydraulic state
surface. This latter is defined by Van Genuchten [7] andassumption of exponential thermal effects done by Gatmiri[3]:

Yd+1 . The yield function ensures the convexity of the elastic

2.3 Transfer laws

(7)

Vvap = - D Tvap ∇(T ) + D Pvap ∇( s )as follows:gg+ g M ε M ij + s ε Svδ ij + T ε Tvδ ij
3
3

1 + +
f d (Ydij , Ω ij ) =Yd 1 Yd 1 −C0 −C1δ ij Ω ji
2 ij ijρw

ψ R (θ w ) = ( Pw − Pa) / γ w

ε M pq1 ∂β (Ω )1 ∂β (Ω )− ε Sv s kl ε Sv − ε Tv T kl ε Tv
2
∂Ωij
∂Ωij
2

ρ vap

Vw and Vvap refer to liquid water and vapour relative
velocities, respectively. ψR is the relative hydraulic potential:

S
Mij
Sv
Tv
ij
The Helmholtz free energy
iswritten as the sum of damaged elastic energies and residualgS
strain potentials [6]. g M Ω ji ε Mji ,δ ij Ω ji ε SV and3gTδ ij Ω ji ε TV are residual strain potentials, which quantify
3
the remaining openings due to cracks after unloading. gM, gS
and gT are rigidity-like scalar material parameters. The
derivation of the free energy ψ S (ε Mij , ε Sv , ε Tv , Ω ij ) ,∂Denmij (Ω kl )(6)

− K w .∇( z )ψ S (ε Mij , ε Sv , ε Tv , Ω ij ) =ψ (εψ R (θW ) dσ (T )1 σ (T )K w .∇(T ) +K w .∇( s)σ (Tref ) dT
γ w σ (Tref )

s≥0

(11)

s0

Only the intrinsic water permeability, which depends on the
behaviour of the solid skeleton, is influenced by damage [6]:K int ij (n, Ω rs ) = k w0 10αw e δ ij + k2ij (n frac , Ω ij )
rev

(12)

The crack related component k 2ij (n frac , Ω ij ) has been
computed by introducing an internal length parameter. The
704

same approach has been used to model the influence of
cracking on vapour flow [4]Air and heat flows have been modeled by using the same flow
rules as an intact porous medium [1]. Damage affects flows
indirectly and isotropically, through the damaged porosity[4].
3. PARAMETRIC STUDY
Here we are interested in the effects of different Van
Genuchten water retention curves introducing by A to E in
Fig. 2 , on the variation of different parameters of damage
model.
The numerical simulation has been inspired from thelaboratory test of Pintado (2002) [8], and the simulation has
been performed with θ-Stock Finite Element code [1]. The
objective of the Pintado test is to apply a controlled flux ofheat at the top of a cylindrical bentonite sample, the bottom
being maintained at a constant temperature (Fig. 3 and 4). The
initial pore water pressure (pw0) is calculated by inversion of
the function of the degree of saturation state surface. The
initial saturation degree Sw0 is equal to 0.63 like in theexperiment conditions. After a heating period of one week, arelaxation period of seven weeks is observed. All of theimposed boundary conditions are given on Fig. 4.

Figure 1. Bentonite heating test [8]. Scheme of experimental
devices.
Table 1. Parameters of the five Van Genuchten water
retention curves
Graph
α(VG)
n (VG)
1.2
A
1.429
B
1.429
C
1.429
D
1.5
E

Figure 2. Five used Van Genuchten water retention curve705

Figure 3. Bentonite heating test [8]. Mesh.

Figure 4. Bentonite heating test ([8]. Boundary conditions.

Figure 5. Space evaluation of suction of specimen B.

As shown in Fig. 5 and 5 in the higher levels of specimens
which are drier since they are near the heat source, themagnitude of suction increases. It reaches to its minimum
value at the bottom of specimens. During the heating period,in the top part of specimen suction goes up while it decreases
in the relaxation period, besides in the lower part of specimen,
suction falls whereas it increases in the remaining sevenweeks. It can be clearly seen that the direction of liquid flow706

is from the top of specimen to its bottom; while it inverses

during the relaxation period.

Figure 6. Evaluation of suction of all specimens at the whole testing time

Figure 7. Evaluation of damage force in all specimens at the whole testing time.

707

Figure 8. Space evaluation of damage force in all specimens at time= one week.

Fig. 7 and 8 also show that the part of the sample which is theclose to the heater has higher amount of damage force,
therefore it is the most damaged part. The Fig. 7 shows by
increasing the heating time,

Yd+1 − rr grows, whereas

Yd+1 − rr falls in the relaxation period. Each graph of Fig. 7
compares the amount of damage force in the different Van
Genchtun water retention curves, it can be implied that thedamage force is produced earlier in the drier specimen. Itshould be mentioned that at the top part of specimens thegraphs get wetter from A to E. There is damage force in thelowest element of specimen at beginning times of test while
there is no initial damage force, but it vanishes in theresuming times. This anomaly occurs in numericalsimulations and it is because of the effects of boundary
conditions.
4 CONCLUSION
Fractured unsaturated porous media is studied with the“THHMD” damage model [3], formulated in independent
state variables. A representation of damage is required to
predict the evolution of fracturing in the neighbouring of
excavated galleries. Saturation variations around gallerieshugely influence the Excavation Damaged Zone. That is whydamage has to be included in hydraulic transfer models andmechanical damage theories have to be extended tounsaturated porous media. A parametric study has been
performed on a cylindrical sample according to laboratory
test of Pintado. Distribution of suction and damage force inbentonite sample has been presented and discussed.

.
5 REFERENCES[1] Gatmiri, B., and C. Arson. , “θ-STOCK, a powerful toolof thermohydromechanical behaviour and damage
modeling of unsaturated porous media,” Computers andGeotechnics, 2008, 35(6): 890-915[2] D.G. Fredlund, N.R. Morgenstern, “ Stress state variables
for unsaturated soils,” J. Geotech. Engng Am. Soc. 1997,
Civ. Engrs;103, 5, 447-466[3] Arson C., & Gatmiri B. , “ On damage modelling in
unsaturated clay rocks,” Physics and Chemistry of the
Earth, 2008, 33:S407-S415.
[4] Arson, C., & B. Gatmiri., “ Parametric study on theperformance of a THM damage model for unsaturated
porous media,” 1st International Symposium on
Computational Geomechanics, 29 April - 1st May 2009,
pp. 553-562, Juan-les-Pins, France[5] Halm & A. Dragon, (1998) , “ An anisotropic model ofdamage and frictional sliding for brittle materials,” Eur. J.
Mech. A/ Solids, 17(3):439-460[6] Arson C. & Gatmiri B. , “ A mixed damage model forunsaturated porous media,” Comptes-Rendus del'Académie des Sciences de Paris, section Mécanique,2009. vol. 337, pp.68–74[7] VanGenuchten, M. , “ A closed-form equation forpredicting the hydraulic conductivity of unsaturated
soils,” Soil Science Society of America Journal 44, 1980,
892-898.
[8] Pintado X., Ledesma A., Lloret A. , “Backanalysis ofthermohydraulic bentonite properties from laboratory
tests,” Engineering Geology, 2002, 64,91–115

708

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Estimation of the Landslide Dam Sustainability in the Gigantic SeimarehLandslide Using the Rate of Sedimentation
Zieaoddin Shoaei
Soil Conservation and Watershed Management Research Institute, Tehran-Iran I.R.
ABSTRACTOn the border of Ilam and Lorestan Provinces in southwest of Iran, and in the heart of Zagros Range of mountains, lies
the gigantic Seimareh Landslide that has been globally recognized as one of the largest mass movements of rocks in theworld. Beside the geomorphic alterations in the landslide site and the depositional area, a major change is thedisturbance of drainage pattern and closure of the rivers. Also, common post-failure feature is landslide lakes which
formed behind the displaced material blocking the rivers. Seimareh and Jidar Landslide dams over Seimareh andKashkan Rivers formed Seimareh and Jidar Lakes. The present study aimed to explain the formation and the time ofwashing off of the landslide dams, using the annual sedimentation rate data and measuring the sediment deposit in theselakes. It was concluded that the Seimareh Landslide dam has been in place for about 935 years after the landslide event.
Keywords Landslide Dam, Lake sediment, Dam Sustainability

1. INTRODUCTION
Aproximately, 50 percent of Iranian land is mountainous.
Due to steep slopes, relatively high precipitation, andvariations in geological settings along with frequent
earthquakes, these vast mountains are prone to slope
instabilities and landslides. Seimareh Landslide in southwestof Iran (Fig.1), is one of the largest known landslides in theworld. The first document reporting Seimareh Landslide waspublished in 1935 during oil exploration projects [1] thatcalled this landslide as a landslip in 1:100,000 prepared
geological map. The most recent study was published by theauthor [2] with a complete review on previous studies andsome detailed field works. Beside the geomorphic alterationsin the landslide site and the depositional area, a major change
is the disturbance of drainage pattern and closure of therivers. Also, common post-failure feature is landslide lakes
which formed behind the displaced material blocking therivers. Seimareh and Jidar Landslide dams over Seimarehand Kashkan Rivers formed Seimareh and Jidar Lakes. The
present study aimed to explain the formation and the time ofwashing off of the landslide dams, using sediment deposit
thickness in these two lakes and the annual sedimentation
rate.
Methodology
In the first step, a detailed plan map was produced on which
the boundaries of the landslide area, displaced debris andlandslide dam lakes sediment were defined (Fig. 1). Then, byusing the topographic map and field data from selectedbenchmark, the longitudinal section along the axis of the
Seimareh River was prepared.
The hydrology of the rivers in the landslide area and some in
the adjacent basins were studied. Assuming that climaticconditions of the landslide area at the time of failure
remained unchanged, the thickness of Jidar and Seimareh

Lakes and the rate of sedimentation in these two lakes werecompared with the suspension sediment load in these rivers atpresent. The result was used to measure the standing period
and the time of wash off of these dams.

Fig. 1 Plan view of detached block, deposited debris,
generated lakes (top) and cross section of Seimareh
Dam and Seimareh Lake along Seimareh River(bottom).
2. Study Area
The study area was selected to cover the entire landsideanatomy, the affected post-failure area, and also thegeological features of the surrounding region. This zone was
located between 32˚: 50́E and 33˚: 15́E, and 47˚: 30́N to 47˚:
30́N. These coordinates cover the mid-Zagros Range with709

NW-SE trend that forms the mountainous regions of thesouthwestern provinces of Iran.
The study area is part of the Zagros range including a seriesof smooth and mild folding with NW to SE trend along thefault system of the area. The presence of gypsum andgypsum lime with severe erodibility resulted to deep valleys
formation in the area. Thus, a relatively wide valley is formed
between northern and southern walls. The Siemareh valley
was formed between Kabirkuh and Mallehkuh (Fig. 1). Mt.
Kabirkuh is emerged as limestone ridge anticline of
Mid-Cretaceous in the southern part of the area.

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

2T

P

2T4

2T4

2T4

2T4

2T4

2T4

2T4

2T4

R

River

Water

Sediment

km2

m3 /sec

x109 m3/yr

g/ litt /m3

x106 t/yr

Kashkan

9400

43.2

1.344

6.00

0.006

8.064

Seimareh

28400

125.4

3.900

1.60

0.0016

6.245

Total

38800

168.6

5.244

7.60

0.007614.309

P

2T4

2T4

2T4

2T4

R

Table 1. Kashkan and Seimareh Rivers discharge and thecorresponding sediment yield

P

P

P

P

P

P

P

P

P

P

2T4

2T

P

AnnualSediment yield

2T4

P

P

Discharge

2T4

2T4

P

Total
Volume of
Water

2T4

2T4

P

P

P

2T

P

Discharge

PP

3. Sustainability of Seimareh Dam
The Jidar Lake, as the second largest one, has an area of about
46km2. It has been formed by the debris blocking Kashkan
River (Fig. 1). The maximum possible height of the Jidar
Dam (which is washed away) is 50m at the southern part andcould hold approximately 1.15 billion m3 of water. Thus,
considering Kashkan River water discharge (1.344 billion
m3/yr), (Table 1) it is expected that the dam was overflowedby the second year after the occurrence of landslide. Anotherlargest post-failure feature in the region is Seimareh Lakethat is completely drained and dried.

Area of the Basin

3.1 Seimareh Landslide
Among dozens of landslides observed on aerial photos andsatellite images in Zagros range, Seimareh Landslide is
identified as the largest one in the entire region. The complex
and mega landslide of Seimareh is one of the largest
instability events not only in this region, but also among allcontinental landslides all over the globe. The massive
Seimareh Landslide in west of Iran is located at the boundaryof Ilam and Lorestan provinces (33˚: 03́ N and 47˚: 41́ E).
Seimareh Landslide is in a basin south of Pole-Dokhtar City
(Fig. 1). This plain is surrounded by Mt. Malehkuh in thenorth and Mt. Kabirkuh in the south. Kashkan and SeimarehRivers are currently flowing in this plain and are called
Karkheh River after merging and continuing towards
Khuzestan Province in south-west of Iran. The megalandslide of Seimareh has occurred in southern part of thisplain and northern hill slope of Mt. Kabirkuh. SeimarehLandslide is initiated by a block sliding of Cretaceous
limestone and its underlying sediment with 16km length,
6km width, and average thickness of 400m, detached fromthe edge of Mt. Kabirkuh (3200m). Subsequently, a 38km3 ofdebris flowed over two valleys in south of Kabirkuh to a
distance of approximately 20km on an average slope angle ofabout 5˚. Along the way, the slope angle of the route
becomes -5˚ (upward) at some areas.
Assuming that the average density of the hard limestone,
shale, and sandstone is 2.1 t/m3, the displaced material might
be estimated at over 79 billion metric tons, qualifying the
Seimareh Landslide as one of the few mega landslides on theearth [2], [3]. The detached and displaced material has
covered an area of approximately 200km2 with a mixture of
mainly limestone and its underlying formation such as marland limestone with low sorting ranges from fine grains tolarge blocks. Due to the extremely uneven topography andundeveloped drainage network in the area, many ponds withvarious areas still exist in the debris of the landslide (Fig.1)
Remnants of the thick lake sediments in the area indicate
the existence of a huge lake subsequently formed behind thelandslide dam. The remains of buried vegetation at thebottom of the lake sediments can be used to date the landslide
occurrence. Oberlande [4], Watson and Wright [5], and van
Zeist and Bottema [6] conducted some research on dating thelandslide by means of C 14 isotope from the Seimareh lake
sediments. Reviewing the available literature and resources
and comparing their results showed that the approximate dateof Seimareh Landslide is around 10000 (±120) years BP.

3.2 Landslide Dam Lakes
Beside the geomorphic alterations in the landslide site andthe depositional area, a major change is the disturbance ofdrainage pattern and closure of the rivers. Because of the lackof developed drainage network, many ponds formed on thedeposited material. Currently, some of these ponds are
utilized as water supply sources for locals. Another common
post-failure feature is landslide lakes which formed behindthe displaced material blocking the rivers. In Seimareh
region, due to the enormous volume of displaced material,
these features were extremely vast. Seimareh Landslide dam
over Seimareh River formed Seimareh Lake (Fig. 1).
Western portion of the debris clogged Kashkan River to form
Jidar Lake on the southern part of Pole-Dokhtar city. At theeastern part of the debris, Balmak Lake was formed likely byblocking some temporary streams in the north of Chenareh
anticline. Amongst these three lakes, Seimareh and Jidar
Lakes were drained naturally and dried out later.

2T4

Figure 1 shows that Seimareh Dam must have had 100 mheight over Karkheh River resulting in a dam with a reservoirof 46.3km in length and a total area of 259km2 after being
filled up. By using the measured altitude of the selectedbench marks and other field surveys data, it seems that thedam reservoir should have contained 11.84 billion m3 ofwater.
Considering the discharge of Seimareh River (Table 1),
Seimareh Lake must have overflowed 4 years after thelandslide occurrence. However, thick lake bed sediments in

710

P

P

P

P

these two lakes indicate that the dams have not been abruptly
washed away and there has been plenty of time for sedimentaccumulation in the lakes. Thus, measuring water yield andthe volume of the lakes is not a consistent dating method fordam standing period.
The entire sediments of these two lakes (Seimareh and Jidar)
are persevered at some spots and can be used as a better
evidence to estimate the sustainability of the landslide dams.
Lake sediment structures such as cross beddings,
laminations, graded-beddings, convoluted structures, andnodules, and also biological structures including root-casts,
plant remains, and freshwater gastropod shells appear inthese sediments beds that make its boundary distinct from thesurrounding sediments. Because of high discharge of
Kashkan River and smaller reservoir of Jidar Lake, it ispossible that, after some certain time, Jidar Lake has drained
into Seimareh Lake. To find out the average thickness of thelakes sediments a thorough field survey was conducted.
Although because of the ragged bedrock it was not easy toestimate the sediments, thus, by dividing the area into several
blocks and using sections of each block, the average sediment
thickness of 60m in Seimareh Lake and 30m for Jidar Lakewere estimated. Sediment beds become shallow towards the
upstream.
The basic concept using the sediment thickness for dating is
illustrated in Fig. 2. This model was employed to estimate thestability period for Jidar and Seimareh Lakes. At the time oflandslide occurrence (T 0 ), sedimentation began in Jidar andSeimareh Lakes at the rate for Kashkan River and SeimarehRiver sediment yield separately. This continued until T 1 ,when Jidar dam broke and the total sedimentation in
Seimareh Lake presumably became the sums of Kashkan and
Seimareh Rivers. Finally, at T 2 , Seimareh Dam broke tobring the sedimentation process to its end. In this study, toestimate the stability of the landslide dams (the time period
between landslide event and the break of the Seimareh Dam)a rate of annual sediment yield over the total sediment wasapplied.

Fig. 2 A conceptual model for sedimentation process anddam stability

by applying the field density of the sediments, the weight wasconverted to volume and then to the thickness of sedimentbeds. The estimated thickness was compared with themeasured sediment beds thickness in the field to evaluate theperiod when Seimareh Dam has been standing. Because Jidar
Dam was washed away earlier with a major impact on
Seimareh Dam sedimentation process, the time of standing
should be measured separately. Applying the data presented
in Table 1, the following measurements were carried out:d Jy =(QK )× (86400 × 360)× (S K )(γ d )× ( AJ )d Jy =(43.2) × (86400 × 360) × (0.006) ≅ 0.093m(1.7 ) × (46 × 106 )Where, d Jy is the thickness of sediment for each year, Q k isdischarge of Kashkan River per second, S k is sediment yieldof Kashkan River (sediment per unit volume of water), γ d isdensity of sediments at Jidar Lake, and A J is the area coveredby Jidar Lake sediment.
Then, by dividing the total thickness of the sediments
measured in Jidar Lake by the annual sediment thickness
increment, the time required for sedimentation, or the period
when Jidar Dam was standing, could be calculated asfollows:

TJ =

30
dJ
=
≅ 325 yrd Jy 0.093

Where, T J is the period of Jidar Dam standing, d J is thesediment thickness measured in Jidar Lake, and d Jy is theannual sediment thickness increment.
During this period, the sedimentation continued in Jidar Lakeand Seimareh Lake received the sediment carried by
Seimareh River:d S1 =

TJ × S S
γ d × ASd S1 =

325 × 6,245,000≅ 5m
1.7 × 259 × 106()

Where d S1 is the sediment deposited in Seimareh Lake before
the break down of Jidar Lake, S S is weight of annual sedimentyield of Seimareh River, ), γ d is the density of sediments, andA S is the area covered by Seimareh Lake sediment.
As the 5m of sediments in Seimareh Lake was formed
when Jidar Dam was standing, then:Based on the hydrographs of 30 years sediment yield records
at Pole-Dokhtar and Karkheh River stations, the weight of theannual sediment carried by these rivers was estimated. Then,711

d S 2 = d S − d S1
d S 2 = 60 − 5 = 55m

Where, d S is the total thickness of sediment beds in Seimareh
Lake at present, d S1 is previously defined, and d S2 is the
thickness of sediment deposited in Seimareh Lake after the
break of Jidar Dam.
To estimate the time required for 55 m of sedimentation in
the Seimareh Lake, the same approach as used for Jidar wasapplied:d K +S =(QK + S ) × (86400 × 360) × (S K + S )(γ d ) × ( AS )d K +S =(168.6)× (86400 × 360)× (0.0076) = 0.0905m(1.7 )× (259 ×106 )Where d K+S is the annual sedimentation in Seimareh Lakefrom Kashkan and Seimareh Rivers, Q K+S is total dischargeof the two rivers, S K+S is the two rivers sediment yield perunit volume of water, γ d is the density of lake sediments, andA S is the area covered by Seimareh Lake sediments
Finally, considering the sedimentation rate and the thicknessof deposited sediment after the break down of Jidar Dam, thetime requires for 55 m of sedimentation can be calculated:

TS =
TS =

dS2
d K +S

55≅ 610 yr
0.0905

Where T S is the time requires for 55 m of sedimentation, dS2
is the thickness of sediment deposited in Seimareh Lake
before the break of Jidar Dam, and d K+S is the annual
sedimentation in Seimareh Lake from Kashkan and SeimarehRivers' sediment.
The total time for Seimareh Dam standing is the sum of times
taken for 30m sedimentation in Jidar Lake and 55m in
Seimareh Lake:-Thus, for accurate estimation of the period of Seimareh dam
stability, investigation of the lake bed sediments seemed to
be a better approach. Comparing the sediments entering
Seimareh Lake (based on the recent 30-year data) with thevolume and thickness of the deposited sediments, theperiod of Seimareh Landslide dam standing was calculated
to be 935 years, which seems to be a reasonable estimation.
-For more accurate dating of landslide occurrence andbreaking of the Seimareh Dam, more reliable dating
methods such as cosmic ray dating and more precise carbon
dating is recommended.
Acknowledgment
This study is a part of an International Program on Landslide
IPL (IPL M-134) and was supported and financed by Soil
Conservation and Watershed Management Research Institute(SCWMRI), Ministry of Jihad-e-Agriculture, Iran.
References:[1] Harrison, J. V., N. L. Falcon and P. B. Mailing 1935, The
Geology of Lorestan from areconnaissance, NIOC report
No. 490.
[2] Shoaei, Z. and J. Ghayoumian, Seimareh, the largest
complex slides in the world In Proc. 8th Int. Cong. on
Engineering and the Environment, 1998, pp. 1337-1342.
[3] Shoaei, Z., Seimareh Landslide, western Iran; one of theworld's largest complex landslides, In Landslide News
Letter, 2000, Vol. 13, pp.23-27[4] Oberlander, T., The Zagross Stream: a New Interpretationof Transverse Drainage in an Organic Zone, 1965,
Syracuse Univ. Press, Syracuse Geography Ser. 168
pages.
[5] Watson, R. A. and Jr. H. E. Wright 1969. The Saidmarreh
Landslide, Iran. Geol. Soc. Am. Inc., Sp. pp. 123,
115-1387.
[6] van Zeist, W. & S. Bottema, Palynological Investigation inWestern Iran, 1977, Palaeohistoria, XIX. 20-85.

T = TJ + TS
T = 325 + 610 = 935 yr

Sustainability of Seimareh Dam for 935 years after thelandslide event might be due to its composition of debris thatformed a resistant structure.
Conclusions-The study of post-failure events can be a good indicator to
study and clarifying the history of the landslide dam andpost events.
-Assuming that climatic conditions of the landslide area at
the time of failure remained unchanged, the recent 30-yearhydrometric data indicate that the Kashkan and SeimarehRivers could fill the largest lake of Seimareh in two or three
years. However, because of the permeability of the dam
body or some overflow from the banks of the dam, thefilling time estimation may not be very accurate.
712

Influence of soil and rock mineralogy and geomorphology on landslides Occurrence and soil erosion in the
Merek Catchment, IranMosayeb Heshmati*a, Nik Muhamad Majid**b, J. Shamshuddin***c, Arifin Abdu**d and Muhamad Ghaituri**
Department of Watershed Management, Agriculture and Natural Research Center, Kermanshah, Iran.
**
Department of Forest Management, Faculty of Forestry, Universiti Putra Malaysia, 43400 UPM, Serdang,Selangor, Malaysia***
Department of Land Management, Faculty of Agriculture, Universiti Putra Malaysia, 43400 UPM, Serdang,Selangor, Malaysia
a. heshmati46@gmail.comb.

nik@putra.upm.edu.my; fax: +603-89432514

c.

samsudin@agri.upm.edu.my; fax: +603-89434419

d.

arifinabdu@putra.upm.edu.my; fax: +603-89432514

Abstract
Part of the mountainous semi-arid areas of Iran is geologically and topographically susceptible to landslides andgully erosion. A study was conducted at the Merek catchment, Iran which is geologically characterized by limestone,sandstone, shale and marls so as to determine the soil and mineralogical factors responsible for the occurrence oflandslides and erosion. In this study, geomorphological units within the catchment were mapped by overlapping
geological, topographical, land use and erosion data, using satellite image and GIS. Rocks and soils were sampled
within each unit to determine their physico-chemical and mineralogical properties. The respective affected areas oflandslides, gully, rill and inter-rill was found 16, 41, 18 and 24%, being eroded at the rate of 10–18.5 t ha-1 yr-1.
These landslides and gullies were initiated by crack formation in areas having smectite and smectite/mica,
respectively. Smectite can swell and contract, depending on the availability of water and was found the dominant
mineral in about 80% of the area affected by landslides. Field verification showed that landslides occur mainly in theforest where the canopy cover is less than 15% with slope of 10-45% and triggered through crack and piping
phenomena due to deforestation and overgrazing, whereas gullies mainly are dominated in agricultural area where
slope is gentle (less than 10%) and developed by agricultural activities, especially improper tillage practices in thiscatchment.
Key words: Geology; Gully: Landslide; Marl; Piping; Rill: Smectite
1. Introduction
Landslide and gully erosion occur frequently in the semi-arid regions of Iran. They are contribute to damage to theenvironment worldwide through soil disturbance and sedimentation. These phenomena are reported as the dominantsoil degradation process in the steep slope and agricultural areas, respectively and affected by improper humanactivities such as road construction, land use change, tillage practice and natural plant clearance [18], [25], [26], [22],[23]. However, gully and landslide are considerable in the forest and arable lands in the Zagros Mountain of Irancomprising fine grained sedimentary rocks (marl and shale layers) that produce soils with high erodibility [7]. The
economic loss due to landslides in Iran is significant, with the estimated annual fiscal loss of about USD 600 million

713

[17] and their off-site impacts such as siltation, soil nutrient decline and CO 2 emission are also significant. In the
study area, suspended sediments through landslide and gully contribute to reduction in incapacity of Karkheh hydroelectrical dam which has been built in down stream. These soil erosion features are accelerated through humanactivities such as forest clearing, overgrazing and improper tillage practices resulting in carbon emission. In the
deforested areas, landslide incidence contributes to transferring 0.94 ± 0.23 Mg C ha−1yr-1 to aquatic ecosystem [9],[4], [23].
Marls in the Zagros are susceptible to landslides where are devoid of vegetative and slope is 20-40%, while there
were prone to gully erosion with gentle slope and agricultural activities [2]. Weathering of marls (containing mica)usually results in the formation of smectite. In some areas where marl layers are overlain by limestone, swelling ofsmectite causes sliding of soil layers. This soil mineral shrinks upon dehydration that leads to cracking phenomenon
because of its low shear strength and can expand up to 30% of its volume [14], [8], [5]. Smectite increase weight ofsoil layer during the rainy season due to high water absorption capacity and consequently sliding incidence. Without
doubt landslide and gully has increased over the years partly due to uncontrolled forest clearing, mining, road
construction and unsuitable agricultural activities. The
objectives of this study were (i) to determine the importantfactors affecting landslides and gully occurrence in semi-(a)arid regions in the Merek catchment, Iran; and (ii) tocompare these factors among different agro-ecologicalzones.
2. Materials and methods
2.1. The study area
This study was conducted at the Merek catchment, locatedin Kermanshah, Iran. It is the upper catchment of the
Karkheh River Basin (KRB) within the Zagros Mountain
Chains (western Iran) (Fig. 1). This catchment has an area(b)of about 23,000 ha that lies between 34º 00' 38" and 34º 09'
3" N and 47º 04' 25" and 47º 22' 18" E, comprising plains,
hilly and mountainous areas with forest, rangeland andagricultural lands. The mean annual precipitation andtemperature are 481 mm and 17.7 °C, respectively.
2.2. The geomorphological units
A major part of the Merek catchment is occupied by the
sedimentary rocks which have been folded into a series ofhuge anticline and syncline [13]. Marl, limestone, sandstone714

Figure 1: a- A map showing the location of
the Merek catchment, b- Geomorphologicalunits with incidence of landslide and gully
erosionand shale are the important materials/sediments forming the geological formations. Geomorphological unit (facies) isa homogenous area in the catchment with specific characteristics of geology, topography, land use and erosion for
management practices and soil conservation measures [1]. This map was prepared by overlapping the digitized mapsof geology, topography (slope steepness, elevation), erosion features and land use type using satellite image (Landsat ETM+ 2002) and GIS software (ILWIS 3.5).
2.4. Rock/soil sampling and analyses
Rock/soil samplings and field verifications were carried out within each geomorphological facies. About 270 soil
samples at the depth of 0-20 and 20-40 cm were taken followed by stratified random soil sampling and theircoordinates were determined by GPS. Soil physico-chemical characteristics were determined in the laboratory using
standard methods for particle-size distribution, aggregate stability, pH, organic carbon, EC, Exchangeable sodium
and carbonate content and their results statistically were analyzed using SAS software. The mineralogy of the soil
and rock (XRD) was determined according to the standard procedure [26].
3. Results
3.1. The geology
Field verifications and geology map showed that geological formations are mainly prone to landslides and gully
hazard. The north part of study area hilly topography consists limestone with intermittent layers of shale andconglomerate resulted in limited occurrence of surface landslides and gully erosion (Gurpi and Ilam Formations),
while in the southern part outcrop of Kashkan Formation shows the presence of claystone, siltstone and sandstone,
which are more subjected to landslide in the steep slope and its deposits (Quaternary) in the gentle slope comprisematerials containing marl, argillaceous materials, sand and silt resulting in gully erosion.
3.2. Topography and land use
As shown in Fig. 2, the respective areas of landslides occur are forest, rangeland and agriculture. In contrast, gully
erosion is occurs mainly in the agriculture area where soil at the Quaternary has been inherited mainly as same aslandslide areas (Kaskan Formation). Although soil, geology and mineralogical factor affecting to landslide incidence
roughly is similar to gully erosion, topographic and land use characteristics are different between them. Field
verification showed that landslides occur at 10-60% slope and is more frequent where slope is 20-45% with15001700 m above the sea level, while gully is dominant erosion feature in gentle slope (<10%) with 1400-1500 m above
the sea level. Landslide in Iran occurred frequently where the slope was 20-40% [1]. In the present study, it wasobserved that most of the landslides occur in the northern aspect of the mountain chain. Soils in the northern slope
aspect can maintain more moisture than others due to the low evaporation [20]. This condition can provide water forthe expanding smectite. In addition, main land use of landslide is forest, but for gully is agriculture. The depth of thelandslides in arable land was found to be shallower than that of the forest and rangeland zones because of its lower
slope. In reality, about 3400 ha (15%) of the study area is affected by landslides (mainly forest zone), but gully
affects 9367 ha (about 41%) which are roughly arable lands.
3.3. The physico-chemical properties of the soils

715

Table 1 shows the soil physicochemical properties. The soils in the landslide and gully areas are mainly clayey innature. The respective average sand, silt and clay were 17, 42 and 41 percent. T-test of particle-size distribution
showed that there was no significant difference within soil layers in all zones, but the ANOVA analysis revealed thatsand value in the agriculture zone is significantly lower than that of the other zones. Inter correlation analysis also
explored that clay particle was correlated with sand, SOC and pH. Clay particles control soil strength and shearing
mechanism when their proportion is more than 30% and water holding capacity is strongly controlled by the clay
content that can be increased up to 210 mm water/m soil depth [8] and [10]. On the other hand, increase in saturation
percentage enhances water absorption capacity that increases the weight of the slipping layer in the slope area orcontributes to gully process through washing away fine fractions.
Although the mean aggregate stability (AS) of the surface soil layers was about 59% in the study areas, there wassignificantly lower in the agriculture areas promoting piping and gully erosion. There was a good inverse relationshipbetween soil aggregate stability and piping density on clayey soils [6]. By and large, the contribution of the physicalcharacteristics on the incidence of landslides in Merek catchment was found to be more than those of the soil
chemical properties due to low values of
ESP, Na and EC as well as moderate
alkaline pH (Table 1). The respective pH,

Table 1: ANOVA analysis of soil variables in the agro-ecologicalzones of Merek catchment, Iran
Soil Variable

Soil
Depth
(cm)

kg-1 and 0.70 dS/m indicating the moderate

Sand (%)

alkaline soil and not expected to have any

Silt (%)

significant relationship with the landslides

Clay (%)

0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40
0 - 2020 - 40

exchangeablesodiumandelectricalconductivity values were 7.75, 0.50 cmolc

and gully erosion. The mean organic carbon
level in the topsoil and subsoil of the
agriculture, rangeland and forest was 1.35,
1.56 and 2.14%, respectively and t-test and
ANOVA analyses showed that their means
are significantly different from each otheramong agro-ecological zones. The burningof crop residues and improper tillage
practices, especially in the hilly areas of the
agriculture zone has reduced

its OCsignificantly. Improper tillage practices

Saturation
percentage
Aggregate
Stability (%)
pH
Exchangeable sodium
cmolc kg-1
Electrical conductivity(dS/m)
Organic carbon(%)
Carbonate (%)

Agro-ecological Zone

Agricult
ure
16.46 b*16.50 b42.80 a42.88 a39.84 a41.13 a51.31 a49.59 a53.41 b57.42 a7.72a
7.81a
0.444 ab0.490 a0.768 a0.711a1.36 a1.03 b
32.18 a36.94 abRangeland
19.33 a19.55 a39.62 ab41.64 ab40.55 a42.20 a50.86 a505 a60.95 ab62.25 a7.76 a7.85 a0.412 b0.408 b0.632 a0.58 b
1.56 b
1.35 a30.01a33.51aForest19.50 a19.33 a37.63 b37.68 b42.70 a41.30 a51.07 a51.14 a63.89 a63.94a7.80a
7.84 a0.512 a0.517 a0.742 a0.778 a2.15c
1.45 a32.18 a40.31b

Pr> F

0.007
0.021
0.003
0.048
0.284
0.866
0.044
0.631
0.0151
0.248
0.195
0.479
0.102
0.021
0.02210.004
0.00010.00010.388
0.018*Means with the same letters are not significantly different at P≤ 0.05% using SNK testfor variables

strongly accelerate soil erosion processes by destroying soil organic matter and soil structures [19]. Overgrazing
contributes to the reduction of OC through defoliation phenomenon and improper tillage practice can decrease about
12% of soil organic carbon [11], [24]. Heavy sheep grazing decreased about 16.5 kg of OC per ha [15]. Carbonate
content in the soils of the study area is high (about 35%) and carbonate indirectly affects incidence of landslides

716through the formation of hardpan, especially in some areas of the rangeland. This in the end would increase theincidence of shallow landslides.
3.4. The mineralogy of rock and soil
Fig. 1 (a) gives the XRD diffractogram of the clay minerals present in the gully areas showing the presence of peaks
at14.04Å, 7.213 Å, 3.301 Å, 3.030 Å, and 2.104 Å indicating
the presence of vermiculite, kaolinite, quartz, calcite
and dolomite, respectively. Soils in this area contain
about 40% clay which potentially contributes to gully
erosion, but there was 12.060Å (K), 9.960 Å (Mg) and10.140Å (Mg-glycol) indicating the presence of micasmectite. Fig.1 (b) is the Mg-treated sample in thelandslide areas showing a peak at 14.81 Å, indicating
the presence of smectite, vermiculite, chlorite or all ofthem. Glycolating the Mg-saturated sample produced a(b)peak at 16.184 Å (Fig. not shown). This shows clearly
the presence of smectite in the sample. Treatment with
K gave a peak at 12.755 Å, while peak at 14 Å peak
was absent (Fig. not shown). This shows that chlorite isabsent in the sample. Field observations revealed that

Figure 1: X-ray diffraction patterns of the clay
fractions treated with Mg; a)-from gully erosion areas(agriculture) which is dominated by vermiculite, andb)- dominated by smectite in the sliding areas (forest),

crack formation was the first step in the development ofboth landslide and gully. The cracks are formed in thefine grained layers containing smectite, mica-smectite

and vermiculite, running parallel and across the direction of the slope. The soils having these minerals can expand
when absorb water, while during dry season, they loss their and consequently the crack formation with different sizes(1-10 cm). Smectite can absorb water (up to 30% of its original volume) resulting in swelling of the crystals andincrease in weight during rainy season resulting in instability and sliding [14], [12], [16], [3].
5. Conclusions
Landslide and gully are an important feature of land degradation in the Merek catchment, Iran that has affected about
56% of the area. Their occurrence is somewhat related to the high soil erosion intensity, which was estimated about
18 t ha-1 yr-1. Incidence of landslides and gully are mainly influenced by the geological, topographical and soil
P

P

P

P

physical properties. Landslides occur frequently on the geological formations, which are composed of fine grainedmaterials, developed from marl and shale strata. These layers weathered to form smectite and vermiculite, which are
subjected to the process of swelling and shrinking, depending on the availability of water. These phenomena
decrease soil stability and eventually result in the occurrence of soil erosion and landslides. The situation is madeworse by the uncontrolled deforestation, improper tillage practices and overgrazing. Most of these landslides andgullies are initiated by cracking and piping processes. Cracks appear in the soils when smectite and vermiculite loss

717

their water during dry season; however, during wet season the mineral absorbs water and expands. This phenomenon
accelerates soil erosion in the Merek catchment, which in turn, increases cracking, piping and landslides/gully. Field
verification showed that both landslides and gully erosion are mainly accelerated by forest clearance and up-down to
the slope tillage practice and landslide was more frequent where plant canopy of forest was less than 20 percent.
Acknowledgements
The authors would like to acknowledge Universiti Putra Malaysia (UPM) and Agricultural Research and Education
Organization of Iran (AREO) for financial and technical support.
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Ardal Region of Chaharmahale Bakhtiari province, Iran. Natural Resources (Iran), 54: 323-329.
[3] Azañón, J.M., Azor, A., Yesares, J., Tsige, M., Mateos, RM., Nieto, F., Delgado, J.,López-Chicano, M., Martín,W., Rodríguez-Fernández, J, 2010. Regional-scale high-plasticity clay-bearing formation as controlling
factor on landslides in Southeast Spain. Journal of Geomorphology, 120(1-2): 26-37.
[4] Blanco, H., Lal, R. , 2008. Principles of Soil Conservation and Management. Springer publisher, USA.
[5] Borchardt, G., 1989. Smectites. In: J.M. Bighman, Dixon, J.B., Milford, M.H., Roth, C.B., Weed, S.B., (Eds.),
Minerals in Soil Environments ( 2nd edition) Soil Science Society of America, Madison Wisconsin,USA,
pp. 967-728.
[6] Diaz, R.P., Sanleandro M.A., Sanchez F.B., Faulkner H., 2006. The causes of piping in a set of abandoned
agricultural terraces in southeast Spain. CATENA, 17: 233-245.
[7] Eberhardt, E.K., Luginbuehl, M., 2005. Slope instability mechanisms in dipping inter-bedded conglomerates andweathered marls in the 1999 Rufi landslide, Switzerland. Engineering Geology, 77: 35-56.
[8] Frydman, S.T., alesnick, M., Geffen, S., Shvarzman, A., 2007. Landslides and residual strength in marls profiles
in Israel. Engineering Geology, 89: 36-46.
[9] Glade, T., 2003. Landslide occurrence as a response to land use change: a review of evidence from New Zealand.
CATENA, 51(3-4): 297-314.
[10] Hazelton, P., Morphy, B., 2007. Interpreting Soil Test Results. CSIRO Press, Sydney,
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[11] Hodgkinson, K.C., 1993. Tactical grazing can help maintain stability of semi-arid woodlands 17th International
Grassland Congress, Palmerston North, New Zealand, pp. 75-76.
[12] Istanblluoglu, E., Bras RL., Cervantes, H.F., 2005. Implication of bank failures and fluvial erosion for gully
development: field observations and modeline. Geophysical Research Earth Surface, 145: 50-70.
[13] Karimibavandpoor, A., Hajihosaini, A., Shahandi, M., 1999. Geological Map of Kermanshah, Iran (1:100,000,
Series: 5458). Geological Survey Service Press, Tehran, Iran.
[14] Krenz, J., Lee, B., Owens, P., 2000. Swelling Clays and Septic Systems: High Water Tables and Septic System
Perimeter Drains. Department of Agriculture (USDA), USA. www.ces.purdue.edu/extmedia/RW/RW-1W.pdf.
[15] Li, C., Hao, X., Zhao, M., Han, G.W., 2008. Influence of historic sheep grazing on vegetation and soil
properties of a desert steppe in Inner Mongolia. Agriculture, Ecosystems and Environment, 128 109-116.
[16] Lutengger, A.J., Cerato, A.J., 2008. Surface Area and Engineering Properties of Fine-grained Soil. Universitiyof Massachusetts Press, Amherst, USA.
[17] Jaada, M. 2009. Landslides Hazard Analysis Using Frequency Ratio Model, Universiti Putra Malaysia, Serdang,
pp.181.
[18] Morgan, R.P.C., 2005. Soil Erosion and Conservation. Blackwell Publisher, Oxford, London[19] Rosa, D.L., Romero, M.A., Pereira, E.D., Heredia, N., Shahbazi, F., 2009. Soil specific agro-ecologicalstrategies for sustainable land use - A case study byusing MicroLEIS DSS in Sevilla Province (Spain). Land
Use Policy, 26: 1055-1065.

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[20] Ohlmacher, G.C., 2000. The relationship between geology and landslide hazards of Atchison, Kansas andVicinity. Kansas Geological Survey, Current Research in Earth Science, Bulletin 244, part 3.
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[21] Page, M., Trustrum, N., Brackley, H., Baisden, T., 2004. Erosion-related soil carbon fluxes in a pastoral steepland catchment. Ecosystems and Environment (New Zealand), 103: 561-579.
[22] Parkner, T., M., M. Page, N. Marden. and Marutani, T. 2007. Gully systems under undisturbed indigenous
forest, East Coast Region. J. Geomorphology (New Zealand), 84: 241-253.
[23] Preuth, T., Glade, T., Demoulin, A., 2010. Stability analysis of a human-influenced landslide in eastern
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Characterization and Classification in New Zealand Soil Bureau, NZ Soil Bureau Scientific Report 79, 27.

719

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The effect of historical land use on landslide initiation
Gholamreza Shoaei
Assistant Professor, Engineering Geology Group, Geology Section, Basic Sciences Faculty, Tarbiat Modares
University, P.O. Box: 14115-175, Tehran, Iran; shoaei@modares.ac.ir

ABSTRACTAn unlikely combination of relatively low rainfall intensity and gentle slopes produced numerous landslides and debris
ﬂows during a storm in July, 2006, around Okaya City, Japan. A detailed investigation was conducted on alandslide/debris ﬂow near Okaya City, where the highly variable soil characteristics led to examine the variability ofphysical and hydrological properties throughout the soil proﬁle. A zone of black organic soil found at depths ranging
from 20 to 280 cm upslope of the landslide scar is indicative of sequential forest burning practiced during Jomon age
hunting and gathering. Previous C-14 dates indicate that anthropogenic practices took place between 10,200 and 4510
years BP at this site. The results of field surveys, laboratory tests, and computer simulations show that the existence ofan ancient anthropic organic soil at the middle of the soil profile altered the hydraulic response of the soil, and could be
the cause of the landslide. Removing the organic ancient soil made the slope stable and thus, the landslide might not be
initiated.
Keywords: Anthropic Soil, Historical land use, Soil profile variability, Rainfall induced landslides, HydrologicalModeling
1. INTRODUCTION
Landslides are important natural geomorphic agents thatshape mountainous areas and redistribute sediment in gentler
terrain. Much of the present Earth’s landscape has been
extensively sculpted by episodic landslides [1]. Landslides
depend on a combination of variables such as slope, rainfall
amount, rainfall intensity, soil mantle thickness, soil stateparameters, subsurface hydrology and vegetation [2], [3].
Reference [4] showed that probabilistic analysis which
ignores the spatial variability of undrained shear strength oreffective cohesion of the soil underestimates the probabilityof slope failure. One of the major factors affecting slope
stability is land use pattern. Most of the tremendous numberof researches dedicated to unveil such effect have focused on
the relatively recent impact of land use change (e.g.
population growth, timber harvesting, forests clear cut, road
construction) within the last few decades [5], [6], and [7].
Their approaches and consideration of land use change, only
through past century, might be reasonable and acceptable;
however, in this research, a new consideration on the possible
effect of ancient land use is raised and could be a new
approach for understanding the cause of some slope failures.
An unlikely combination of relatively low rainfall intensity
and gently sloping topography produced numerous landslides
and debris flows during a prolonged storm in July 2006
around Okaya City, Nagano Prefecture, Japan, and indicatedthe existence of other factors contributing to the landslide
initiation. This research focused on detailed investigation on arelatively large landslide/debris flow near Okaya City (Fig. 1)
where the highly variable soil characteristics in the landslide

scarp led to examine the variability of physical andhydrological properties throughout the soil profile and within
the area above the landslide.

Fig.1 The location of study area
Based on 62 yr of rainfall data from the nearest AMeDAS
meteorological station in Suwa City (6 km east of thelandslide site), July (during the Baiu season) has the highest
total monthly rainfall. Among all 62 yr, July 2006 had thehighest rainfall for this month (506 mm). Average July
precipitation near the study area is 200.5 mm. Total
precipitation during the July 17-19, 2006 period exceeded
358 mm. Landslides were triggered when cumulative 3-day
rainfall reached about 330 mm, most of the total July rainfall.
However, the storm had a relatively low intensity; maximum
intensity was 22.5 mm h-1 near the beginning of the storm andit never exceeded 16 mm h-1 throughout the next 3 days.
The region is located approximately 53 km to the east ofJapan's second largest volcano, Ontake. The geology of thearea is relatively simple because of its volcanic nature andalso due to its location in the middle of the main island of720

Japan (Honshu). Several local faults exist in this region;
however, it seems that the geological setting has not beensignificantly affected. Dominant lithology in the area is tuff,
andesite, and pyroclastics. Andesite breccias containing large
amounts of tuff breccia composed the lower layer ofPleistocene bedrock.
Soils around the landslide scarp are about 3.5 m deep andoverly andesitic bedrock. Soils at this landslide site areformed primarily by erosion processes, where transported
soils are deposited over approximately 0.5 m of residual soil.
Soil profiles at the site indicate considerable small-scale
heterogeneity and anisotropic characteristics. For example, asequence of black soil layers appears near the middle of the
profile, whereas the soil overlying the bedrock below theblack layers is brownish. The anthropic black soil in the study
area is an ancient buried surface soil remnant from earlier
land use (e.g., episodic burning of vegetation), which waslater covered by eroded soil. The spatial propagation of thisblack soil layer appears discontinuous. The soil around thelandslide is dominated by the black layer, whereas a nearby
landslide behind the same hill contains brownish soil overbedrock overlain by a thin layer of organic-rich soil on thesurface (i.e., no black layer). Repetitive agricultural activities
at the study site formed more than a meter of such organicrich surface soil [8].
2. METHODOLOGY
To investigate the role of soil profile heterogeneity, a detailed
investigation on soil variability and its effect on pore pressure
distribution through and after the storm was planned
including:I) An integrated and precise study of the soil including
vertical and spatial characteristics and distribution, physical,
and hydrological properties; variation in texture andstructure; and composition and genesis of the soil. Soilcharacteristic curves and Ks of the soil profile were measurein the laboratory for different soil layers. In addition, an
extensive study was conducted on the history of the region to
discover the stages of land use around the study area. This
was to figure out a comprehensible scenario for soil
heterogeneous profile forming process.
II) Field measurements from multi-depth tensiometers tomonitor soil water potential (both tension and positive porewater pressure) fluctuations during recorded rainfall tocapture the real-time response of the soil to the storms. To
model the water movement and development ofsuction/pressure fluctuations in the soil due to rain, eighttensiometers were installed in field above the investigatedlandslide. The tensiometers were arranged in two sets, each atfour different depths: 20 cm, 50 cm, 100 cm, and 167 cm.
This depth arrangement allowed the tensiometers to monitor
the middle of the surface soil, middle of subsurface soil,
bottom of the subsurface soil, and a deeper soil at the middleof the profile, respectively. To reach the soil below the black
layer and monitor the effect of this low permeable soil layer,
the deepest tensiometers was installed at 167 cm below theshallow black layer. A group of four tensiometers was set up a
few meters above the landslide. The second set was installed

10 m further upslope from the first set. An eight-channel
logger connected to the transducer and powered by a carbattery (for longer life) recorded soil suction data at 10 min
intervals.
III) Modeling the process of rain water infiltration into thesoil by applying field data from different soil horizons,
including inherent variations. The model will be calibrated
based on the recorded tensiometer data and simulations arethen being conducted for the landslide-inducing storm of July
2006 to capture the process of transient unsaturated-saturated
flow in the soil. The goal of this study was to simulate theinfiltration and propagation of rain water within the soil, bothvertically and laterally in an at least 2-D model, which caused
the landslide of July 19, 2006. Thus, a suitable model needs tobe selected to fulfill some of the vital criteria for simulation.
These model criteria can be summarized as: 1) ability toincorporate transient rainfall; 2) handle large pressure orwater content gradients between the elements of the slope
allowing accretion of sharp pressure head fronts progressing
through the soil during short-term, high intensity rains; 3)
model soil layering effects in the profile with variablehydrological properties; 4) work with an acceptable numberof soil properties; and 5) numerically solve the governing
equation so it can be easily compared and fitted to fieldobservations. Considering all the parameters mentioned
above, HYDRUS 2D/3D (Pc-Progress Co.) seems to fulfill
most of the requirements for the modeling in this study by:offering a user friendly interface in modeling preparations
steps, being capable of a large iteration process and time step
control to optimize the required running time for thesimulation, receiving the topographic and soil layer data froma prepared file, and thus accepting incredible slope details,
and giving a good control on ‘mesh elements’ generatingprocess. Thus, HYDRUS 2D/3D was chosen to conduct themodeling.
3. RESULTS AND DISCUSSION
The study area, located west of a large fresh water lake anddrained by several streams, was an appropriate location forsuch habitation and agriculture starting from the early Jomon
period 14,000 to 4000 BC (e.g. [9]). The black soil layer is anindicative soil horizon around this study area near Okaya City
which supports the existence and propagation of Jomon
culture throughout the region. 14 C dating analysis of charcoal
in the black layer indicates a wide age range from 10,200 to
4510 yr BP [10]. The beginning of Jomon era coincided witha warmer climate in Japan [9]. This warmer climate supported
flourishing primitive practices, such as hunting andassociated burning of forests around Okaya City for an
extended period. The Jomon people lived by food gatheringand hunting, and therefore this destructive interaction withnature began. There is no firm evidence indicating the processof hunting, whether it was a group activity or an individual
practice, however, it involved hunting and clearing the forests
and regenerating vegetation by fire, possibly on an annual
basis (e.g., [9]). The anthropogenic origin of the black soil has
been mentioned in some previous publications (e.g. [11]),although it has never been connected chronologically to any
particular era or ancient civilization. Based on 14C dating721

results, burnt wood and brush accumulated over the soil forabout 5700 yr until about 2500 BC. More results anddiscussion about the genesis of the soil and its possible
historical origin are presented in reference [8].
Data collected on soil thickness and distribution by fieldsurveys show that the soil mantle at the area above thelandslide can be divided into two portions: I) shallow soil
with maximum depth of 1 m (Fig.2), and II) deep soil from 1
to ≈4 m deep.

The scenarios of the modeling were: I) when the anthropic
black layers exist in the soil profile (the actual case); and II)
Fig.4.theCalibrated
on tensiomers
datathewhen
black layersmodelof soilbased
are removed
to simulate
recorded
on
July
2007.
effect of human activities in the study area. The model was
run for the storm in July, 17-19, 2006, as 88-h of rain (Fig 5).

Fig.5. The storm of July, 17-19, 2006.
Fig.2. Location of shallow and deep soil mantle
The delayed initiation of the landslide in the July 2006
indicates that cumulative rain exerts a greater influence on
critical pore pressure accretion than rainfall intensity. Thisphenomenon is important related to simulating the porepressure response in the deeper soil zone, since it captures theeffect of delayed hydrological response of the soil, due towater percolation process within the thick soil layer. To
control the simulation of soil response to the storm, six
observation points (in the model, Fig. 3) was set to read thepore pressure at the landslide spot and on particular depths.

The result for the first set of simulations is presented in Fig 6.
The curves represent the pose water pressure at theobservation nodes 0.20 m to 3.50 m at the bottom the soil on
the bedrock contact. The fig shows that at the time oflandslide initiation and around the soil-bedrock contact, porewater pressure exceeds 3.00 mH 2 O.

Fig.6. The model simulation for actual soil profile.

Fig 3. The modeled soil and slope and observation nodes.
Fig 4 shows the model simulations for after the longprocess of calibration at the depths of tensiometers
installation. As shown, the simulated pore pressures fit thetensiometers response. After this stage, the model was set torun for the storm of July, 2006, to simulate the pore pressure
at the site of the landslide.

On the other set of runs, black anthropogenic soil layers were
removed to simulate the propagation of pore water pressure in
the absence of human effects. Fig 7 presents the result of porepressure in the lack of black soil layer. The pore waterpressure does not rise above 2.50 mH 2 O when there is noanthropogenic black layer. It seems that the black soilintercept the lateral flow of the water from upslope to build ahigher pore pressure around the time of failure, whereas, inthe absence of the black soil layer, the water passes throughthe slope to build a delayed pore pressure at the location oflandslide. Such difference in the built up pore water pressure
may change the timing of the failure. Considering that the
failure occurred toward the end of storm, the increase the in722

pore water pressure might not continue to make the slope
unstable and cause the landslide.

Fig. 7. The model simulation for No-Black soil scenario
4. CONCLUSION
In this study, a 2D model was applied to the entire
contributing area above the actual landslide. This model
included all the details of the slope characteristics and soil
properties in order to examine the effect of theanthropogenic soil historically embedded into the soil
profile. The simulations showed the effects of soil layering
as a result of historical land use and anthropogenic activities,
and thus, provide a unique look into the extended influenceof human activities. Large differences were found between
the hydrological and stability simulation results with original
slope conditions and he assumed scenarios of no humaneffects. The more accurate simulations that were observedwith anthropogenic effects included in the model suggestthat the historical human life styles and land use activities
affected contemporary mass movement phenomenon by: 1)
being the probable cause of the 2006 failure, and 2) altering
the timing of the landslide by changing the response ofpressure head distribution in the soil during the storm, which
likely could change the magnitude of economic and humanlife loss in such events (July, 2006, occurred early morning).
Thus, the result of this study may introduce a new concept toassess the historical aspects of anthropogenic soil oncontemporary landslide behavior. In contrast to earlier
studies, this research extends the concept of land use and its
effect of landsliding from conventionally a few decades tohistorical land use. Because such a long history affects thegenesis, and thus the layering and properties of the soil, andimposes heterogeneity and anisotropy, the effects ofhistorical anthropogenic activities have a direct link to thecontemporary physical, hydrological, and mechanicalcharacteristics of the soil. The results suggest that although
recent human activities may affect the frequency and thenumber of landslides, the ancient anthropogenic activities
play a more chronic role by gradually contributing tocatastrophic mass movement events over time.

6. REFERENCES[1] Sidle, R.C., Ochiai, H., Landslides Processes, Prediction,and Land Use, American Geophysical Union, Water
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DC, USA.
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management. In: Wignosta, M.S., Burges, S.J. (Eds.),
Land Use and Watersheds: Human Influence on
Hydrology and Geomorphology in Urban and Forest
Areas, Water Science and Application. American
Geophysical Union, 2006, Washington, p. 195–227.
[3] Rybar, J., Stemberk, J., Wagner, P., Landslides:
Proceedings of the First European Conference on
Landslides, Prague, 2002, Czech Republic, Taylor andFrancis, London.
[4] Griffiths DV, Fenton GA. Probabilistic slope stability
analysis by finite elements. J Geotech Geoenviron Eng
ASCE, 2004;130(5):507–18.
[5] Sidle, R.C., Ziegler, A.D., Negishi, J.N., Abdul Rahim,N., Siew, R, Turkelboom, F., Erosion processes in steep
terrain – Truths, myths, and uncertainties related to forest
management in Southeast Asia. Forest Ecology andManagement, 2006, 224(1-2):199-225.
[6] Anderson, M.G., Holcombe, Liz, Renaud, Jean-Philippe.
Assessing slope stability in unplanned settlements indeveloping countries. Journal of Environmental
Management, 2007, 85:101–111[7] Imaizumi, F., Sidle, R.C., Kamei, R., Effects of forest
harvesting on occurrence of landslides and debris flows
in steep terrain of central Japan. Earth Surface Processes
and Landforms, 2008, 33:827-840.
[8] Shoaei, G., Sidle, R.C., Variation in soil characteristics
and hydrologic properties associated with historic land
use near a recent landslide, Nagano Prefecture, Japan.
Geoderma, 2009, 153:37–51.
[9] Yoneda, M., Suzuki, R., Shibata, Y., Morita, M.,
Sukegawa, T., Shigehara, N., Akazawa, T., Isotopic
evidence of inland-water fishing by a Jomon population
excavated from the Boji site, Nagano, Japan. Journal ofArchaeological Science, 2004, 31:97–107.
[10] Okamoto, T., Katakura, M., Togashi, H., Shimizu, Y.,
Kitahar, H., Ochiai, H., Kawasaki, N., Kume, Y., Ages ofburied humus horizons in landslide disaster sites causedby July 2006 heavy rainfall around the Suwa Region,
Nagano Prefecture, Proceedings of the General Meetingof the Association of Japanese Geographers, 2007,
72:133 (in Japanese).
[11] Yamanoi, T., Geological investigation on the origin of
the black soil, distributed in Japan, Journal of GeologicalSociety of Japan, 1996, 102(6):526-544; (Abstract inEnglish)5. ACKNOWLEDMENTThis paper is a part of an extensive study on July 2006
landslide. I am grateful to Dr. Roy C. Sidle and Dr.
Ken’ichiro Kosugi for their helps and inputs through thisstudy at Kyoto University. Also special thank to
Monbokagakusho for funding this study.
723

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051 77

Integrated Environmental Management for Sustained Development
J.RAJARAMAN 1 and K.THIRUVENKATASAMY2AMET UNIVERSITY CHENNAI INDIA

ABSTRACT
Lowland problems are many. Many urban settlements are on the coastal belt, worldwide. In the past, many urban areas
grew in an uncontrolled manner. The urban populations experienced increased risks, a poorer quality of life, and healthproblems. However, in general, there has been increasing management of many urban environments through Land use
planning, building, environmental, health and safety regulations and management of emergencies, collectively
balancing social, economic and environmental requirements. In this paper the complexities of urban cities are illustrated
through many examples without naming any particular city. Since in modern times, globalization has created anenvironment where economic borders only exist. The systems are usually administered by separate authorities, which
need to act together to secure sustainable urban management. Therefore actions and responses are frequently
disconnected and are sometimes conflicting. To attain sustainable growth common sources of information arenecessary. To emphasis the following eleven illustrative examples of some urban cities and their issues are discussed.
Keywords: Environmental Geology, Aquifer ,Sustained development, Geomorphology.
1. INTRODUCTION
Urban areas are extreme cases where human activities havethe greatest impacts on the local environment, andenvironmental processes can affect the greatest numbers ofpeople. They need separate attention. Decisions made by
developers and urban authorities have major implications forhealth, safety, the economy and the environment. Since
ground processes and the nature of the underlying materials
influence the occurrences of geological resources and hazards,
it is important that these should be properly understood sothat appropriate actions are taken to safeguard natural
resources, reduce risk to people.
In this paper the need for good communication between urban
authorities , developers and geoscientists.
2. GROWTH AND RENEWAL IN URBAN AREAS
The geographical origins of towns and cities are diverse.
Most develop in response to natural and local conditions. The
following factors influence settlements:
1. Convenient stopping places on routes such as passes
between hills, convenient points on rivers, or at water
sources in arid areas.
2. Ports on sea coasts, rivers or lakes.
3. Location in farming areas where routes meet allowing
the development of markets.
4. Defensive sites, often on rocky prominences with areas
below that could be used for building dwellings.
5. Administrative centres normally at prosperous
settlements developed for other reasons but sometimes
separately, for instance to refocus activity by removing
administration from a coastal town to a site in theinterior to promote development.

724

All but the last of the above mentioned factors are associatedwith specific Geomorphological settings that reflect
underlying geology
3. MANAGEMENT OF THE URBAN ENVIRONMENTIn the past, in many countries, especially in developing
countries urban areas grew in an uncontrolled manner. The
urban population experienced increased risks, a poorer
quality of life, and increased health problems. This trend
continues where migration from rural to urban areas is very
rapid. However, there has been increasing management ofmany urban environments through a variety of regulatory
regimes.
1. Land use planning – which is concerned essentially withlocating development in the most appropriate places, while
protecting areas worthy of conservation, so thatdevelopment is undertaken at the best balance of social,economic and environmental cost.
2. Building regulations- which aim to ensure that built
structures are properly constructed and are fit for use.
3. Environmental regulations- regulating the operation andoccupancy of sites, ensuring that waste and emissions areproperly controlled and dealt with and that air and water areprotected, so as to minimize social and environmental
damage, including risks to health, conserving natural
habitats and maintaining biodiversity .
4. Health and safety regulations – to ensure that
employees, and other people entering places of
employment, are safeguarded from unnecessary risks.
5. Environmental health regulation – to ensure that people
are not exposed to potential hazards to health.

6. Management of emergencies – setting out anddisseminating procedures for protection, evacuation andrelief, making provision for emergencies services inrelatively low hazards locations, and minimizing the
likelihood of major accidents.

4.3 THE CITY SCENARIO 3
A major manufacturing city was linked to a sea port by anaging railway that had been constructed through a range ofhills, along a river valley. Because of poor maintenance andincrease in traffic the uncontrolled outcome is shown in fig.3.

4. EXAMPLES OF URBAN ISSUES
4.1 THE CITY SCENARIO 1
In a city an industrial centre grew on the basis of local coaland ironstone mining that supported heavy industry. The
uncontrolled outcome is shown in figure 1.

Fig.3

Fig.1
4.2 THE CITY SCENARIO 2
In the initial stage small quarries were opened for the
extraction of limestone many years ago. The stones were usedin the construction of major local buildings. The quarries
were then abandoned. The buildings became part of thecultural heritage. The city expanded over the stone resourcearea. The uncontrolled outcome is shown in fig.2.

4.4 THE CITY SCENARIO 4
A city is located beside a bay with a narrow coastal plain
backed by mountains. It was a port but expanded along thecoast as housing, industry and tourism developed. Because ofdredging the uncontrolled outcome is shown in fig.4

Fig.4

Fig.2

725

4.5 THE CITY SCENARIO 5
A city had been a major manufacturing centre. More recently,
traditional industries had declined and a thriving service
sector was now the main employer. The excess ground water
created uncontrolled outcome as shown in fig.5

4.7 THE CITY SCENARIO 7
Porous sandstone beneath part of a city is the major source of
local water suppl. The aquifer is overlying in places by sandand gravel and elsewhere by low permeability clay. A major
chemical works contributed uncontrolled outcome as shownin fig.7.

Fig.7

Fig.5
4.6 THE CITY SCENARIO 6
A major industrial area developed in and around a city. The
complex was powered by coal and by manufacturing of gas
and coke. Ash was tipped on areas between the factories andgave rise to localized contamination. This contamination
caused uncontrolled outcome as shown in fig.6

4.8 THE CITY SCENARIO 8
An area of uplands was rich in timber of high commercial
value. Because of the need to earn foreign currency thegovernment licensed major logging concessions. The logging
company wished to maximize its profits and undertook rapid
tree felling over a few years. Cutting of trees createduncontrolled outcome as shown in fig.8

Fig.8

Fig.6

726

4.9 THE CITY SCENARIO 9
The low lying coastal area, lined by dunes, marshes and areasof mangroves was located to the west of a major port that had
been designated for expansion. This involved the building oaf
a breakwater to the east of the undeveloped coast and thedredging of a deeper access channel. The breakwater and thedredging operations created uncontrolled outcome as shownin the fig.9.

4.11 THE CITY SCENARIO 11
An important town grew several hundred years ago on theslopes of the mountain. It was initially a market center forfarming based on rich local soils and plentiful rainfall on theslope. From time to time, there were volcanic eruptions but
most of these were limited to small scale ash falls. But the
volcanic eruptions were very active over the past few hundredyears. The scientists found that many of these cities key
emergency services and a key evacuation route for the
population were located in the area that was at greatest risk.
The volcanic eruptions, seismicity and ground movements
contributed to uncontrolled outcome as shown in fig.11.

Fig.9
4.10 THE CITY SCENARIO 10
Major population growth took place in a city before mains
drainage was constructed severe pollution of surface water
and wells by sewage occurred. The unplanned expansion andpopulation explosion created uncontrolled outcome as shownin fig.10.

Fig.10727

Fig.11
Many more examples could be cited throughout the world.
These examples make clear the diversity of issues that may
affect urban areas and the nature of the information that isneeded to address them. The information comes from many
sub-disciplines, some of which are rooted in Geoscience but
others extend, to Ecology, Economics, Sociology,
Architecture, and Archaeology. Multi-disciplinary studies areneeded. In some cases communication may be straightforward but often there are barriers of unfamiliarity andtechnical complexity that can stand in the way of prompt andeffective action. To achieve satisfactory results it is necessary
to understand the implications of the Geology and Earthsurface processes of the area and to explain them clearly interms that can be understood widely. This needs adequate,
easily accessible, and readily understood information, advice,guidance and training on how to deal with these issues.

5. LOWLAND CITIES AND THEIR ISSUES IN A
COMPREHENSIVE TABLE
Table I Lowland cities and their issues

The solution is “be wise at least after the event.” [1],[2],[3].
6. INFORMATION, GUIDANCE AND
COMMUNICATIONThe eleven city examples discussed bring together the needfor coordination by authorities responsible for land use
planning, environmental protection including catchment
management, health and safety at work, and environmental
health. In some cases land use and emergency planning are
required to solve the problem. In some special cases shorelinemanagement is also needed to regulate the activities.
In many cases the information necessary comes from many
disciplines but main focus is on the Geosciences. Therefore
the following are particularly relevant:

728

1). Resources – minerals, water, soils, landscape, goodbuilding land- types, extent, quality and significance(safeguarding and potential)
2) Hazards and constrains – subsidence, slope instability,
flooding, storms, natural and manmade contamination,
erosion, volcanic activity, seismicity- types, extent andrelative significance. Information is needed on allenvironmental issues that are relevant to a particular area atboth the regional and more detailed levels.
It is important that adequate basic Geological,
Hydrogeological,Geomorphological, Geotechnical isundertaken at the strategic and more local level in order thatapplied outputs and modeling procedures can be securely
based.

7. CONCLUSIONS
1. Traditionally Geosciences information has been presented
on paper maps. These maps are essentially in two
dimensional representations that can be interpreted in three
dimensions by specialist users but not readily by
non-specialists. Environmental and engineering Geology
Mapping will help to overcome the barriers with somesuccess.
2. The urban area has often been regarded as if it intrudes
within and is therefore separate from the Eco-System. But inpractice urban areas have major effects on their surroundings
and processes in nearby areas have impacts on towns andcities. Therefore the urban area should be regarded as anintegral part of the overall functioning of the Eco-System if
environmental issues are to be properly analyzed andresolved.
3. There is a considerable diversity of environmental issues
that have a Geosciences element and many problems involve
multi-disciplinary consideration within a variety ofmanagement systems, often administered by different
authorities.
4. These authorities often need to draw on the same body ofknowledge with varying degrees of emphasis but they can
only do so if they are aware of the relevant issues andsignificance and availability of the necessary information.
5. In practice, there is commonly a communication gap
between varies groups of administrators and specialists, aswell as with the general public and politicians.
6. This arises partly from limitations on education in the
Geosciences and a lack of dialogue and mutual learning .
7. A key aspect is for stakeholders to know who to ask, whento ask, and what to ask when problems arise.
8. A key to sound urban management is the compilation andmaintenance of data bases not only for the urban area but also
for its regional surroundings. These are particularly valuable
if linked to GIS, Environmental information systems andsoftware for modeling potential outcomes of possible
decisions.
9. Experience shows that authorities are often unwilling toinvest in research even though the outlay can lead toconsiderable savings in the future.
10. In addition it is often difficult to secure funds formaintenance of the data system . Few public authorities areequipped to do this even though the cost are very small
compared with the development cost and the bill for putting
problems right after the event.

729

8. REFERENCES[1]. Marker,B.R.(1996).Urban development – Identifying
Opportunities and dealing with problems. In.:McCall
et al (Eds) pp181-214.
[2]. Brook D. and Marker B.R. (1998). Geomorphological
Information neede for Environmental policy formulation
In: Hook,J.M. (Ed) Geomorphology in Environmental
policy J. Wiley (Chichester) pp 247-262.
[3]. Igor S. Zektser Geology and Eco-systems , Springer(2006) pp 123 - 147

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Lowland Environmental Geotechnology of Seismosediments of Kandla Port in
India
J.RAJARAMAN1 and K.THIRUVEKATASAMY2


ABSTRACTThe Gujarat coast in India is vulnerable to various natural disasters such as earthquakes, cyclones, tsunamis andagricultural droughts. The nearby main town on main land is Gandhidham.Kandla is located at a distance of about 4 km
from Gandhidham. A 20 m wide road connects Gandhidham and Kandla. Geographically, Kandla could not be
considered as a port because, there is water on almost 340 0 of this port. Kandla is almost an island. There is only one
escape route via the bridge. In case the bridge gets damaged due to natural disasters, then there is no alternate route forrescue, evacuation and escape. At present huge amount of dredging is regularly being undertaken at the Kandla port. On
an average about half million cubic meters of dredged materials is excavated, carried and is dumped in deep sea. In thispaper an integrated approach to the clay behavior (including seismosediments) is considered mainly to relate cohesion
and friction under different environmental conditions in lowlands, to shear strength by different types of clay.
Keywords: Lowlands,seismosediments, lateral earth pressure, clay minerals
Keywords: Four or Five Keywords
1. INTRODUCTION
In flood affected lowland areas the loss of stability due todrastic variation in clay behavior and lateral pressure is aproblem that needs separate treatment. The accumulated
sediments due to huge amount of dredging adds more
complications to the already existing problems. On an
average about half million cubic meters of dredged materials
is excavated, carried and is dumped in deep sea. The
heterogeneity of the sediments is due to many factors inwhich seismosediments are one of the important factors.
1.1 Seismosedimentation And Seismosediments
Geological materials which are crushed, fractured, loosenedor displaced directly or indirectly as a result of a earthquake
induced processes such as landslides, landslips, slumps,
mudflows, avalanches, liquefaction, rockfalls, rockslides,
tsunamis and soils moved under gravity in the coastal areas orin the vicinity of the river are defined as seismosediments andthe process is defined as seismosedimentation.
1.2 Textural Inversions
Textural inversions also cause mixing of different clay
sediments to create more heterogeneity in soils. For example
quartz Arenites are more poorly sorted and many contain highpercentage of sub angular and angular grains. Some quartz
Arenites exhibit textural inversions such as a combination ofpoor sorting and high rounding, a lack of correlation between
roundness and size, such as small round grains and larger
angular grains, or mixtures of rounded and angular grains
within the same size fraction. These textural inversions
probably result from mixing of grains from different sources,
erosion of older sandstones, or environmental variables such730

as floods, tsunamis and wind transport of rounded grains into
a quiet water environment. Angular grains may result also
from diagenetic development of secondary overgrowth.
1.3 Clay Minerals and Sediments
The dredged material in Kandla port consists of fine sandy
clay or clayey sand. In these sediments mainly three types ofclays are involved. They are Kaolinite, Illite andMontmorillonite. Due to the mixing factors discussed above
the soil becomes highly heterogeneous. The Geotechnicalproperties vary from pocket to pocket in an erratic manner. A
separate treatment of Geotechnical methods for coastal
lowlands is necessary.
2. PROPERTIES OF CLAYS
Plastic property of clay is due to adsorbed water thatsurrounds the clay particles. This property is affected by typeof clay minerals and their proportional amounts in a soil. The
ultimate effect is that different clay soils possess different
liquid and plastic limits.
2.1 Activity of Clay
Skempton defined a quantity called Activity i.e. the slope ofthe line correlating PI (Plasticity Index) and percent finer
than 2μ therefore Activity A

Seed and others studied the plastic property of several
artificially prepared mixtures of Bentanite and Kaolinite. The
results are shown in figure 1

TABLE I Activities Of Clay Minerals

Fig.2
T

[2]

TABLE II
Variation of φ with Types of Clay Minerals

2.3 The Random Nature of Clay

Fig.1

[1]

2.2 Clay Minerals and Angle of Internal Friction φ
For normally consolidated clays cohesion C = 0. The equationof the Mohr – Coulomb failure envelope is given by S = σ'tan φ.Where φ is angle of internal friction and S is shear
strength . Then Sin φ = 1   3.
 1   3(1)

The modified forms of Mohr’s failure envelope for pure clay
minerals are shown in the figure 2. In a consolidated drained
test the total stress is equal to the effective stress, since theexcess pore water pressure is zero. At failure, the maximum
effective principal stress is σ1' = σ1 = σ3 + Δσf, where Δσf, isthe deviator stress at failure. The minimum effective principal
stress is σ'3 = σ3
The modified form of Mohr’s failure envelope of pure clay
minerals and quartz with typical values of φ is grouped as in
Table II

731

Kaolinite and Illite have nonexpansive lattices whilst that ofMontmorrillite is expansive. There is no particular value ofplastic limit that is characteristic of an individual clay mineral
type. Similarly there is no single liquid limit which ischaracteristic of a particular clay mineral. The liquid limits
for Illites fall in the range of 60 – 90 % whilst those for
Kaolites vary from 30 -75%. The presence of 10%
Montmorrillonite in an Illitic or Kaolinitic clay can cause asubstantial increase in their liquid limits. The plasticity
indices of Na and Li Montmorrillonite clays haveexceedingly high values, ranging between 300 to 600 %. The
Mohr-Coulomb envelope only indicates an ultimate shear
strength value. It is the final compromised state ofequilibrium of many random variables.
3. STRESS PATH AND LATERAL EARTH PRESSURE
Consider a normally consolidated clay specimen subjected toa consolidated drained triaxial test fig.3 (a). At any time
during the test, the stress condition in the specimen can be
represented by Mohr’s circle as shown in fig.3(b). In adrained test, total stress is equal to effective stress so σ3 = σ'(minor principle stress).
σ1 = σ3 + Δσ = σ1' (major principal stress).

At failure, Mohr’s circle will touch a line (Mohr-Coulomb
failure envelope). This makes an angle φ with the normal
stress axis.

1. The stress paths for a given normally consolidated clay aregeometrically similar, and
2. When the points representing equal axial strain say (ϵ 1)are joined, will be approximate straight lines passing throughthe origin. Now considering a soil specimen subjected to one
dimensional consolidation type of loading we can write
σ'3 = K0 σ1'(7)This equation (7) is very important because it gives a
relationship between σ'1, σ'3 and the coefficient of earthpressure at rest (K0). According to Jaky, K0 = 1- sin φ. Thisequation connects various types of clay and their respective
angle of internal friction φ. From the following fig. 4 in the
Mohr’s circle the coordinates of point E is given by
(8)
(9)

Fig.3
[3]
Now considering the effective normal and shear stresses on aplane making an angle of 450 with the major principal plane,
Effective normal stress,(2)
Shear stress,(3)
The points on Mohr’s circle having coordinates p' and q' areshown in fig.3b. If the points with p' and q' coordinates of all
the Mohr’s circles are joined, will result in the line AB. Theline is called a stress path. The straight line joining the origin
and point B is defined here as the Kf line. The Kf line makes
an angle ∝ with the normal stress axis. Now

Fig.4
[3]
From the plot of p' versus q' with K0 and Kf lines as shown in
fig.5

(4)
Where σ'1( f ) and σ'3( f ) are the effective major and minor
principal stresses at failure.
Similarly,(5)
From equations (4) and (5) ,tan∝ =sin φ.

(6)
Fig 5

4. SETTLEMENT CALCULATION USING STRESSPATH
From Lambe’s stress path the following conclusions are
made:

732

[3]

The angle(10)

Where β is the angle that the line OE (Ko line) makes with thenormal stress axis.
5. KANDLA PORT ENVIRONMENTThe Kandla Port is surrounded on all sides by water. There
are two creeks viz Nakti creek and the Kandla creek surroundthe port and there is water on about 340 degrees. At the
Kandla end of the bridge, the two creeks almost meet one
another. The above-mentioned route is the only connecting
link between Kandla and Gandhidham.

Fig.7 Sketch of VEF at Kandla[4]

6. THE ENVIRONMENTAL GEOTECHNICALPROBLEMS
The environmental Geotechnical problems are highly
complex. These problems are associated withseismosediments, dredged materials, clay minerals. Basicallythe dredged soils will show different soil characteristics than
the natural soils with natural compaction. Therefore a detailed
micro-studies related to soil mechanics alone will yield goodsolutions. The settlement calculation will clearly indicate thecomplex nature of the problems.
6. 1 Settlement Calculation
An illustrative example is shown in fig.8
Fig 6
[4]
Arun Bapat suggested for a construction of verticalevacuation facility (VEF). Instead of dumping into the deep
sea the dredged material could be used to build a verticalevacuation facility. A vacant and large area of about 100
sq.km is available at Kandla. This land is in S.W. direction
and south of Khari creek. This is very near the port. This land
can be used for VEF.
The facility will be a construction of small hill like structurewith a shape of tapering conical or a flat top cone with one or
two tops. For this purpose, retaining wall along the peripheryof VEF will be required. A hillock of about 15 to 20 m highwith a base of about 200 to 250 m diameter could
accommodate about fifteen to twenty thousand people ormore. The dredged material consists of fine sandy clay or
clayey sand. This could be suitably compacted so that thestructure will be in maximum stability conditions. It could
have a wide ramp with gentle slope for rapid movement oflarge population. This ramp would be winding the small
hillocks in serpentine way. Suitable shady trees could beplanted at various terraces of the VEF so that it becomesstable.Construction of such VEF would help in saving several
thousand lives in the case of tsunami attack at Kandla.
According to the view of Arun Bapat the sketch of the VEF at
Kandla will appear as shown in the fig.7

Fig.8
[5]
Fig.8 represents the stress – strain contours for a given
normally consolidated clay specimen obtained from anaverage depth of a clay layer. Taking compression index
Cc = 0.25 and e0 and the drained friction angle φ as 300 andfrom equation (10),And k0 = 1 – sin φ =1 - sin 300 = 0.5733Knowing the value of β the k0 line is formed as shown infig. 8. Since tan∝ =sinφ, angle φ =300, tan∝ = 0.5.
So ∝ = 26.570. The settlement is calculated for the clay layerfrom the following conditions :1. Insitu average effective over burden pressure
σ'1 = 75kN / m2.
2. Total thickness of clay layer = Ht = 3mOwing to the construction of a structure, the increase of thetotal major and minor principal stresses at an average depth ofΔσ1 = 40kN/m2
Δσ3 = 25kN/m2
Assuming that the load is applied instantaneously. Insitu
minor principal stress at rest pressure is σ3 = σ'3 = k0σ'1 =0.5(75) = 37.5kN/m2.
So before loading,So, the consolidation settlement is
Sc = 0.0087 Ht = 0.0087(3) = 0.0261m and hence the total
settlement is
Se + Sc = 0.09 + 0.0261 = 0.116m.

7. CLAY AND LATERAL PRESSURE
In the following Table III the variation of k0, the lateralpressure and clay type is shown.
TABLE IIIThe stress conditions before loading can now be plotted in
fig.8 from the above values of p' and q'. This point is A. Sinceloading is instantaneous( i.e., undrained), the stressconditions in clay represented by the p' versus q' plot
immediately after loading, will fall on the stress path BCA.
Immediately after loading,
σ1 = 75+40 = 115kN/m2 and σ3 = 37.5+25 = 62.5kN/m2. So,

With this value of q', we can locate point D. At the end of
consolidation, σ'1 = σ1 = 115kN/m2 σ'3 = σ3 = 62.5kN/m2
So,
and q' = 26.25kN/m2.
The preceding values of p' and q' are plotted as point E.
FEG is a geometrically similar stress path drawn through E.
ADE is the effective stress path that a soil element, at average
depth of the clay layer, will follow. AD represents the elastic
settlement, and DE represents the consolidation settlement.
For elastic settlement (stress path A to D),

For consolidation settlement (stress path D to E), based on
previous valid assumption the volume metric strain between
D and E is the same as the volumetric strain between A and H.
Note that H is on the k0 line. For point A, σ'1 = 75kN/m2, andfor point H, σ'1 = 118kN/m2. So the volumetric strain, ϵ v

The axial strain ϵ 1 along a horizontal stress path is about
one-third the volumetric strain along the k0 line, or734

K0 values are different and therefore the settlement anddifferential settlement will be also different. The activity ofeach type of clay (Table I) will further complicate theproblem in the presence of flood water. The clay
Montmorillonite and its swelling potential will be a major
problem.
8. IMPORTANCE OF CLAY IN APPLIEDENVIRONMENTAL GEOTECHNOLOGYInterparticlc friction is the principal property which permits
soils and granular materials to resist load without formation.
Dry soils of virtually all types are highly stable. The suitabilityof soils for engineering purposes depends largely upon theirability to remain in place and support whatever loads may beplaced upon them, either by a permanent engineering
structure or by transient vehicle loads. A study of theproperties which distinguish the more satisfactory from theless satisfactory soils indicates that in the majority of cases
clays are detrimental to stability. It is apparent that wet clay
has the effect of a lubricant in diminishing the naturalresistance due to friction that would otherwise exist. It isnecessary that the civil engineer responsible for constructionof any form of earth work should be informed not only
concerning the Quantity of clay minerals that are present but
also about their nature and their potential influence on theengineering properties of the soil.

As the ability of soils to resist deformation depends verylargely on the internal friction, wet clay has the effect of
reducing or canceling out the frictional resistance. It may also
be pointed out that the so-called cohesive resistance is
induced almost entirely by the clay fractions, and therefore
that clean sands are noncohesive. Again we mustnote the important part played by water, as finely grounddry clay particles exhibit no cohesive properties. Ifwater is added to a dry soil, the cohesive resistance will
normally increase with the addition of moisture and in
most cases the frictional resistance will not be greatly
impaired until a certain amount of moisture is added.
Beyond this point, the friction will diminish but thecohesive resistance may continue to increase up to somepoint of higher moisture content, after which both values
will diminish as the soil approaches a completely fluid
state.
Metals are typical substances having little or no internal
friction; in them, resistance values are almost entirely a resultof the cohesive or tensile strength.
Fig. 9(a) shows wheel load and subgrade in road construction
in lowland coastal area. Fig.9(b) is a chart showing
characteristic curves illustrating loss in stability or internal
resistance of a crushed sandy gravel due to the addition ofincrements of plastic clay.The combined figures 9(a) and 9(b)clearly indicates the instability created by Montmorillonite( Bentonite), Illite (local clay), and Kaolinite in the presenceof water.Bentonite fails first followed by Illite. Kaolinite ismore stable.
9. CONCLUSIONS
1. Every port has its own Geological setup and environmental
Geotechnology.
2. Seismosediments create heterogeneous nature to the soil.
They represent accumulated disturbed samples.
3. The φ value for clay varies with type of clay.
4. Calculation of settlements by using strain methods reveals
more intrinsic nature of clay behavior than stress oriented
methods.

735

5. Increase in stability of a soil is not a function of cohesion
and angle of internal friction alone because water introduces“turn around stages” and create instability especially incoastal area lowlands.
6. In a mixed sediments (seismosediments, flood sediments) a
depth of 3m of clay (say Kaolinite 1m, Illite 1m andMontmorillonite 1m) KIM – will create 24 different
Geotechnical sequences which in turn will increase thecomplexity of the Geotechnical problem. They are :KIM,MIK,IMK,KKK,III,MMM,KII,IKI,IIK,IKK,KIK,KKI,
MII,IMI,IIM,KMM,MKM,MMK,KMI, IMM, MIM, MMI,
MKI, IKM.
7. The final conclusion is that the construction techniques orsite preparation should match with Geotechnical properties ofdifferent mixed types of clays.
10. REFERENCES[1] Seed,H.B.,R.J.Woodward, & R.Lundgren, Clay
Mineralogical Aspects Of The Atterberg Limits, J.Soil
Mech.Found.Eng.Div.,Am. Soc.Civ.Eng.Vol.90,
no.SM4,pp.107- 131, 1964a.
[2] Olson,R.E.,Shearing Strength Of Kaolinite, Illite, And
Montmorillonite, .Geotech.Eng.Div.,Am.Soc.Civ.Eng.
Vol.100,no.GT11,pp.1215-1230,1974.
[3] Lambe,T.W.,Methods Of Estimating Settlement, J. Soil
Mech.Found.Div.,Am.Soc.Civ.Eng.,Vol.90, no. 5,
pp 43- 54,1964[4] Arun Bapat Tsunamigenic Vulnerability of Kandla Portand use of Emergency Vertical Evacuation Facility, Pro. Nat.
Conf. on Coastal Processes,Kerala, pp.378-381[5] Braja M.Das Advanced Soil Mechanics, Taylor andFrancis group Indian reprint 2010
[6] Francis N.Hveem, Importance of Clay in Applied soil
Mechanics ,Clays and Clay Minerals; 1952 v. 1; no. 1;
pp. 191-195;

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Single-Step Extraction to Determine Soluble Lead Levels in Soil
Masahiko KATOH1), Satoshi MASAKI1), Takeshi SATO1)
1) Department of Civil Engineering, Gifu University, Japan
ABSTRACTThis study investigated single-step extraction using EDTA to extract the amount of potentially soluble Pb in soil, which
would be almost as much as that in the first four Pb fractions of Tessier’s sequential extraction procedure (SEP), with orwithout the immobilization amendment hydroxyapatite under different extraction conditions such as concentration andextractant pH. The results clearly showed that 0.1M EDTA (pH 7.5) with a soil/extractant ratio of 1:100 can extract
almost 90% of the first four Pb fractions of SEP, and EDTA-extracted Pb was positively correlated with the first four Pb
fractions even in Pb-immobilized soil. In conclusion, single-step extraction using EDTA was found to be a useful
method to evaluate potentially soluble Pb even in Pb-immobilized soils.
Keywords: EDTA, Immobilization, Pb-Contaminated Soil, Potentially Soluble Pb, Single-Step Extraction

1. INTRODUCTION
Heavy metal contamination, which is common in soil
throughout the world, poses a potential risk to human andanimal health owing to migration through soil profiles into
groundwater. Therefore, various remediation techniques havebeen applied to reduce the mobility and bioavailability ofheavy metals in soil.
To effectively remediate heavy-metal-contaminated soil, it isnecessary to know the amount of toxic elements in the soil.
However, only determining the total concentration of heavy
metals in soil is not sufficient because mobility andbioavailability are strongly dependent on the chemical phasesof heavy metals in soil [1], [2]. Therefore, determining thechemical forms of such heavy metals is especially importantfor chemical remediation such as in-situ immobilization andex-situ soil-washing techniques.
Heavy metals in soil exist in various chemical forms
including exchangeable, inorganic such as carbonates, boundto Fe/Mn oxide and organic matter, and retained in a clay
mineral. Heavy metals bound to Fe/Mn oxide and organic
matter, which are less mobilizable than exchangeable heavy
metals, have the potential to solubilize with changes in soil
conditions such as pH and oxidation-reduction. Therefore, toremediate contaminated soil, especially by using chemical
remediation techniques, it is necessary to evaluate the amountof potentially soluble heavy metals including heavy metals
bound to Fe/Mn oxide and organic matter as well.
A sequential extraction procedure (SEP) has been widelyused to analyze the chemical phases of heavy metals in soil.
Tessier’s SEP divides the chemical phases into five fractions:exchangeable, carbonates, Fe/Mn oxide, organic matter, andresidual [3]. The residual fraction should only be insoluble
under natural environmental conditions, which suggests thatthe first four SEP fractions are possibly being solubilized byion-exchange and sorption/dissolution onto soil particles.
However, SEP is not suitable for analyzing a large number ofdifferent contaminated soils because it is time consuming andthe analytical technique requires special facilities.
Ethylenediaminetetraacetic acid (EDTA) is the most widelyused chelating agent owing to its ability to form strong
chelate complexes with many heavy metals. The extraction
efficiency of EDTA, which is closely dependent on manyfactors of extractant such as its concentration and pH, and thetypes of contaminated soil [4]-[6], has been extensively
studied to predict the mobility and bioavailability of heavy
metals in contaminated soil [7], sediments [8]-[10], andwastes [7], [8] as well as to improve the efficiency ofsoil-washing techniques [6], [11]-[13]. However, only a few
studies have compared the amount of heavy metal extractedwith EDTA and the amount of each fraction of SEP [8]. It has
not been determined whether EDTA can extract the first four
fractions of Tessier’s SEP. Moreover, EDTA extraction has
never been applied to contaminated soil after the addition of
an immobilization amendment that would alter the chemicalphases of heavy metals in soil. To assess the effectiveness ofthe chemical remediation technique applied, it is important tobe able to determine the amount of potentially soluble heavy
metals in soil after chemical remediation.
The present study focused on Pb, which is one of the most
toxic and prevalent heavy metal contaminants in Japan andthroughout the world, and investigated single-step procedureusing EDTA to extract potentially soluble Pb from soil. We
speculated that the amount of potentially soluble Pb would bealmost the same as the amount of Pb extracted in the first four
fractions of Tessier’s SEP. In conjunction with potentiallysoluble Pb, we also investigated whether the same result
could be obtained after an immobilization amendment wasadded.

736

2. MATERIALS AND METHODS
2.1 Study area
This study was conducted at two shooting ranges: Tajimi and
Nakatsugawa located in Gifu, Japan. The mean annual
precipitation and temperature of both shooting ranges are
2200mm, 14°C, respectively and this area lies between 35º
20' 8" and 35º 28' 6" N and 137º 06' 5" and 137º 29' 2" E.
2.2 Soil Preparation and Characterization
The Pb-contaminated soils were collected from depths of
5–15 cm at both shooting ranges. At Tajimi, seven soil
samples were collected at 3-m intervals from the shooting
point where shooter fires a bullet toward the shooting
direction. Table Ⅰ shows the chemical properties of the
Pb-contaminated soils collected at Tajimi (t) and
Nakatsugawa (n). Sample t-1 was collected closest to theshooting point; sample t-7 was collected at a point farthestfrom the shooting point. The Tajimi samples used forchemical analysis were air-dried and passed through a 2-mm
sieve without adding any immobilization amendments. One
soil sample (n-1) was collected at Nakatsugawa.
Hydroxyapatite (hereafter, HAp) was added to sample n-1 at
a ratio of 1, 2.5, and 10% (w/w), hereafter, n-2, n-3, and n-4,
respectively. After adding HAp, water was added to maintaina water holding capacity of 60%, and then the samples wereincubated for 1 week. After incubation, the Nakatsugawa
samples were also air-dried and passed through a 2-mm sieve
prior to chemical analysis.
The water-extracted Pb and total Pb, Fe, Mn, and C contentsof the Tajimi and Nakatsugawa soil samples were determined.
The water-extracted Pb was determined by extraction withwater at soil/water ratio 1:10. To obtain the total Pb, Fe, andMn contents in soil, the soil samples were digested with
HNO 3 and HCl using a microwave oven. All solutions were
filtered with a 0.45-μm filter and analyzed for element
concentrations by ICP-OES (ULTIMA 2, Horiba). The total
C content in the soil was determined by CHO analyzer (JM10,J-SCIENCE LAB). These analyses were not performed on
samples n-2, n-3, and n-4.

2.3 Determination of Lead Phases in Soil by Tessier’s
SEP
SEP was performed on the soil sample following theprocedure described in [3]. 1.0 g of soil that has been passed
through a 0.425-mm sieve was extracted with 25 mL of a 1 M
MgCl 2 solution (exchangeable Pb). The soil remaining after
the first extraction procedure was extracted with 25 mL of a 1
M sodium acetate solution with pH 5 (Pb carbonate). The soil
remaining after the second extraction was further extractedwith 20 mL of 0.04 M NH 2 OH-HCl in 25% (v/v) HOAc in a
95 °C water bath with occasional agitation (Pb bound to
Fe/Mn oxide). The soil remaining after the third extraction
was extracted with 3 mL 0.02 M HNO 3 and 5 mL 30% H 2 O 2solution in an 85 °C water bath with occasional agitation.
After 3 h of extraction, 5 mL 3.2 M NH 4 OAc in 20% (v/v)
HNO 3 was added and the soil was shaken for 20 min (Pbbound to organic matter). The remaining soil was digested
with 5 mL HNO 3 and 2 mL HCl using a microwave oven(residual Pb) and diluted in a 50 mL measuring cylinder.
After digestion was complete, the solution was passed
through a 0.45-μm filter and analyzed for Pb concentration by
ICP-OES.
2.4 Single-Step Extraction using EDTA
The effects of EDTA concentration, extractant pH, andsoil/extractant ratio on the amounts of Pb extracted from soil
were investigated using t-1, t-2, t-3, t-5, and t-7 to evaluate theamount of potentially soluble Pb by single-step extraction(SSE) with EDTA.
2.4.1 Concentration
10 mL of 0.5, 0.1, and 0.05 M EDTA (pH 7.5) was added to a
15 mL polypropylene tube containing 1.0 g soil samples t-1,
t-2, t-3, t-5, and t-7 (1:10 soil/extractant ratio), and then, thetube was shaken for 1 h.
2.4.2 pH
10 mL of 0.1 and 0.05M EDTA adjusted to pH 3.5 using 1M
HCl was added to a tube containing 1.0 g soil (1:10
soil/extractant ratio), and then, the tube was shaken for 1 h.

Table I Chemical properties of Pb-contaminated soils
No.

Sampling location

t-1 Tajimi
t-2 Tajimi
t-3 Tajimi
t-4 Tajimi
t-5 Tajimi
t-6 Tajimi
t-7 Tajimi
n-1 Nakatsugawa
n-2 Nakatsugawa
n-3 Nakatsugawa
n-4 Nakatsugawa*1 Not Analyzed

Amendment

Water-extracted Pb(Type/ratio(w/w))
HAP/1%
HAP/2.5%
HAP/10%(mg/kg)
0.4
0.5
0.7
0.7
1.0
1.3
3.8
20.8
N.A.*1
N.A.
N.A.

737

Pb
(mg/kg)
91
229
1551
986
1341
2060
9289
3966
N.A.
N.A.
N.A.

Total
C
Fe
(g/kg)
8.8
35.8
19
32.6
79
25.8
37
28.3
58
21.0
174 13.8
135 24.8
6.0
30.4
N.A. N.A.
N.A. N.A.
N.A. N.A.

Mn
1.7
1.6
1.2
1.0
1.0
1.0
6.4
6.0
N.A.
N.A.
N.A.

filter and analyzed for Pb concentration by ICP-OES. All
experiments were performed in triplicate, and the averagevalues were calculated.
The extraction ratio in this study, defined as the ratio of
EDTA-extracted Pb to the first four fractions of Tessier’s SEP,
was calculated according to the following equation:

R% =

PbSSE
× 100
PbSEP

(1)

where Pb SSE (mg/kg) is the amount of EDTA-extracted Pb
and Pb SEP (mg/kg) is the amount of the first four fractions ofTessier’s SEP.
EDTA-extracted Pb of the t-4, t-6 samples was determined byextraction with 0.1M EDTA (pH 7.5) at a soil/extractant ratioof 1:100 following the procedure described above.
2.5 Single-Step Extraction using EDTA on Immobilized
Soil
Based on the results obtained in the procedures described in
Section 2.4, the extraction with 0.1M EDTA (pH 7.5) at asoil/extractant ratio of 1:100 would be suitable to predict
potentially soluble Pb in soil. Moreover, the suitability of SSE
for Pb-immobilized soils amended with HAp, which would
gradually alter the amount of the residual Pb fraction in SEP,
was investigated following the procedure described above.

Fig. 1. Sequential extraction of Pb from Tajimi soil

3 RESULT AND DISCUSSION

Fig. 2. Sequential extraction of Pb from Nakatsugawa
soil.
2.4.3 Soil/Extractant Ratio
100 mL of 0.1 and 0.05M EDTA (pH 7.5) was added to a
100mL tube containing 1.0 g soil (1:100 soil/extractant ratio),and then, the tube was shaken for 1 h.
After shaking, all solutions were passed through a 0.45-μm

3.1 Soil Chemical Properties
The water-extracted Pb and total Pb, C, Fe, and Mn contents
in the soil used in this study are shown in Table I. The amountof the water-extracted Pb in Tajimi soil tended to increase
with the distance from the shooting point. The amount of thewater-extracted Pb in both Tajimi and Nakatsugawa soils
exceeded environmental standards for soil contamination in
Japan. The total Pb content in Tajimi soil was the lowest for
t-1 and then increased gradually with the distance from theshooting point. This increase in the total Pb content might be
attributed to the different number of Pb bullets that landed on
the soil. From the evidence that the total Pb content in t-7
exceeded 9000 mg/kg, it was considered that heavy Pb
contamination has occurred in Tajimi soil. The total C
contents in t-6 and t-7 (135–174 mg/kg) were higher thanthose in t-1 to t-5 (8.8-79 mg/kg). The total Fe content ranged
from 13.8 to 35.8 mg/kg. The total Mn content in t-7 was 6.4
mg/kg and was higher than that in the other samples. The
water-extracted Pb and total Pb content in Nakatsugawa soil
were 20.8 and 3966 mg/kg, respectively. The total Pb content
was lower than that in t-7, whereas the water-extracted Pb
was higher. This probably resulted from the difference
between the Pb phases in Tajimi and Nakatsugawa soils. This
result will be discussed in greater detail in the following
section.
From these results, it was confirmed that the soils used in thisstudy had different chemical properties such as

738

Table II EDTA-extracted Pb and extraction ratio under different conditions.
EDTA-extracted Pb
Concentration pH S/E ratio t-1
t-2
t-3
t-5
t-7
(mg/kg)
0.5 M
7.5
1:10
43
175 1029 1096 6552
0.1 M
7.5
1:10
59
203 1214 1554 8747
0.05 M7.5
1:10
68
198 1271 1375 7713
EDTA-extracted Pb (a)
0.1 M
3.5
1:10
68
210 1342 1650 8253
0.05 M3.5
1:10
83
245 1466 1797 8493
0.1 M
7.5 1:100138
365 1130 1100 6796
0.05 M
7.5 1:100141
180
755 1040 7501
*1
156
268 1529 1631 7705
First four fractions of SEP. (b)
0.5 M
7.5
1:10
27
65
67
67
85
0.1 M
7.5
1:10
38
76
79
95
114
0.05 M
7.5
1:10
43
74
83
84
100
0.1 M
3.5
1:10
44
79
88
101
107
Extraction ratio*2
0.05 M
3.5
1:10
53
92
96
110
110
0.1 M
7.5 1:100
88
137
74
67
88
0.05 M
7.5 1:100
90
67
49
64
97
*1: Pb amount of first four fractions of SEP*2: (b)/(a)*100
water-extracted Pb and total Pb, C, Fe, and Mn contents.
3.2 Lead Phases in Soil by Tessier’s SEP
Figs. 1 and 2 show the results of Pb fractions in Tessier’s SEP
for Tajimi and Nakatsugawa soils, respectively. In both
figures, the upper and lower graphs show the amount andratio of the Pb fraction, respectively. The average recovery
ratio, defined in this study as the ratio of the total amountfrom SEP to the total amount of Pb amounts in the soil shownin Table I, was 110±22%.
The ratio of the Pb fraction in Tajimi soil was the highest in
the fraction bound to Fe/Mn oxide and ranged from 42 to59%; the second highest was in carbonate ranged from 15 to
32%. In contrast, the highest ratio of Pb fraction in
Nakatsugawa soil was carbonates; it was around 50% in n-1
with no added immobilization amendments. The amount andratio of exchangeable Pb in n-1 were 460 mg/kg and 12%,
respectively, although those in t-1 were 355mg/kg and 4%,
respectively. In spite of a lower total Pb content in n-1 than in
t-7, the water-extracted Pb in n-1 was higher than that in t-7,
which probably resulted from the greater amount of
exchangeable Pb in n-1 than in t-7.
The residual fractions in n-3 and n-4, where HAp was applied
at 2.5% and 25% (w/w), respectively, were significantly
increased, although the composition of the Pb fraction in n-2,
where HAp was applied at 1% (w/w), was not significantly
different from that in n-1. It is widely known that adding HAp
to Pb-contaminated soil probably results in the precipitationof Pb phosphate minerals such as pyromorphite(Pb 5 (PO 4 ) 3 X: X = OH, Cl, F) by immobilizing Pb in thecontaminated soil according to the following reactions [14],[15]:Ave. S.D. C.V.

62.4
80.4
77.0
83.6
92.2
90.9
73.6

21.3
28.0
21.0
25.0
23.3
27.1
19.8

34.1
34.8
27.3
29.9
25.3
29.8
26.9

Ca 5 (PO 4 ) 3 OH + 6H+ → 5Ca2+ + 3H 2 PO 4 − + OH−(Dissolution of HAp)

(2)

5Pb2+ + 3H 2 PO 4 − + OH− → Pb 5 (PO 4 ) 3 OH + 6H+(Precipitation of pyromorphite) (3)Pyromorphite has a lower solubility product constant at log
K sp = −25.05 [16] and is probably included in the residual
fraction of SEP. From this consideration, pyromorphite might
be precipitated by HAp to increase the residual fraction in n-3
and n-4. On the other hand, the low percentage of HAp added
in n-2 probably caused no difference in the Pb composition
between n-1 and n-2.
From these results, it was confirmed that the soils used in thisstudy had significant differences in the amount andcomposition of Pb phases.
3.3 Single-Step Extraction by EDTA
3.3.1 Concentration
The EDTA-extracted Pb and extraction ratio are shown in
Table II. The extraction ratios for all concentrations tended to
be lower at distances closer to the shooting point. It is widely
accepted that EDTA forms stable complexes with many
heavy metals. Therefore, EDTA can extract Pb sorbed to
Fe/Mn oxide and organic matter in soil as well as inorganic
Pb compounds such as Pb carbonate. However, Palma andMecozzi [10] reported that it was more difficult to extract theexchangeable form compared with other phases. Hence, theextraction ratio in Tajimi soil might be decreased at distances
closer to the shooting point owing to a higher ratio of
exchangeable Pb. In addition, the extraction ratio for 0.5M

739

EDTA was not higher than that for 0.1 and 0.05M EDTA.

Fig. 3. Relationship between first four Pb
fractions of SEP and EDTA-extracted Pb in
Tajimi soil.

samples were 90.9±27.1 and 73.6±19.8% for 0.1 and 0.05M

Fig. 4. Relationship between first four Pb
fractions of SEP and EDTA-extracted Pb
in Nakatsugawa soil.

Table III EDTA-extracted Pb and extraction ratio in Nakatsugawa soil
No.

EDTA-extracted Pb (a)First four fractions of SEP *1 (b)(mg/kg)

n-1
3246
n-2
2039
n-3
1835
n-4
3190
Ave.
S.D.
C.V.
*1: Pb amount of first four fractions of SEP*2: (b)/(a)*100

Extraction ratio*2
(%)
88
68
85
100
85.3
13.2
15.5

3701
2996
2151
3183

According to Elliott and Brown [5], the extraction ratio would
not become greater above a certain concentration. Moreover,Finzgar and Lestan [6] revealed that EDTA extraction
efficiencies decreased as the molar ratio of EDTA to total Pb
in soil increased. From these results, it was considered that
0.1 and 0.05M EDTA should be suitable for the extraction ofpotentially soluble Pb in soil. Therefore, 0.5M EDTA was not
used in the following investigation.
3.3.2 pHSSE using EDTA was performed again with pH changed to
3.5 (Table II). Although compared with that at pH 7.5, theextraction ratio at pH 3.5 was improved for all samples, theratio especially in t-1 was still low at 44% and 53% for 0.1
and 0.05M EDTA (pH 3.5), respectively. To perform asimplified analysis rapidly in the field, it is certainly
recognized that a neutral pH range is suitable for extractant
pH. From these results, it was proposed that an extractant pHof 7.5 would be suitable.
3.3.3 Soil/Extractant Ratio
SSE using EDTA was performed again with thesoil/extractant ratio changed to 1:100 (Table II). Comparedwith the results obtained above, the extraction ratio for 0.1
and 0.05M EDTA (pH 7.5) was improved even in t-1 at 88%
and 90%, respectively. The average extraction ratios of five

EDTA (pH 7.5), respectively, and it appears that 0.1M EDTA
(pH 7.5) with soil/extractant ratio 1:100 may extract
potentially soluble Pb in soil. This extraction condition isconsistent with that for the soil-washing technique proposed
by Garrabrants and Kosson [7].
Fig. 3 shows the relationship between the amounts of 0.1M
EDTA-extracted Pb with pH 7.5 and soil/extractant ratio
1:100 and the amount of the first four Pb fractions of SEP.
The relationship calculated using the software JMP 8 could be
expressed as follow: y = 0.88x − 92 R2 = 0.995***, where y(mg/kg) is the total Pb in the first four fractions of SEP and x(mg/kg) is the EDTA-extracted Pb. The slope of the line was
0.88, suggesting that 0.1M EDTA can extract almost 90% of
the first four Pb fractions of SEP.
The EDTA extraction was performed for the t-4 and t-6
samples under the same conditions, and the results could be
represented by the equation presented above (Fig. 3). These
results indicated that 0.1M EDTA (pH7.5) withsoil/extractant ratio 1:100 can extract almost 90% of the first
four Pb fractions of SEP.
3.4 Single-Step Extraction of Pb from Immobilized Soilwith EDTA
Immobilized Nakatsugawa soil samples were extracted with
0.1M EDTA (pH 7.5) with soil/extractant ratio 1:100 (Table
III). The average extraction ratio was 85.3±13.2% regardless740

of the amount of HAp amended and was approximately thesame as that for Tajimi soil. If the results of the Nakatsugawa
soil samples are plotted in the same manner as shown in Fig. 3,
the amounts of EDTA-extracted Pb can also be represented
by the same equation for the Tajimi soil samples (Fig. 4).
These results suggested that it would be possible to extract
potentially soluble Pb in spite of different kinds of soils andthe addition or non addition of immobilization amendment. In
addition, pyromorphite, which would be precipitated with theaddition of HAp, would remain in soil without dissolution
and/or complexation with EDTA owing to its very low
solubility.
In this study, we have used only two kinds of soils and one
type of amendment. Further investigation should consider
more types of soil and amendment. Moreover, it is desirable
to improve the extraction ratio up to approximately 100%.
However, this study did suggest that it would be possible toextract potentially soluble Pb even in immobilized soil withEDTA.
4 CONCLUSION
This study investigated single-step extraction using EDTA ofpotentially soluble Pb that could be almost as much as theamount of Pb in the first four fractions of Tessier’s SEP. In
conjunction with potentially soluble Pb, we also investigatedwhether the same result could be obtained after the additionof an immobilization amendment. The results obtained in thisstudy were mainly as follows:
1) EDTA extraction under different extraction conditions
from soils having different total Pb amounts and different Pb
phase compositions was investigated. The results clearly
showed that 0.1M EDTA (pH 7.5) with soil/extractant ratio
1:100 can extract almost 90% of the first four Pb fractions of
SEP, and EDTA-extracted Pb was positively correlated withthe first four Pb fractions of SEP.
2) The soils with immobilized Pb using HAp were extractedusing 0.1M EDTA (pH 7.5) with soil/extractant ratio 1:100.
The results showed that the amount of EDTA-extracted Pb
was also almost 90% of the amount of the first four Pb
fractions and could be represented by the same equation asthat of the non amended soils.
3) In conclusion, single-step extraction using EDTA wasfound to be a useful method to evaluate potentially soluble Pb
even in Pb-immobilized soils.
5 ACKNOWLEDGMENT
The ICP-OES and CHO analyzer instruments used forchemical analysis in this study were made available by theDivision of Instrumental Analysis at Gifu University.
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[4] Karamanos R E, Bettany J R, Rennie D A,
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ICP-OES with the HCl and Tessie extraction methods,”
Talanta, vol. 64, 2004, pp. 345-354.
[10] Palma L D, Mecozzi R, “Heavy metals mobilizationfrom harbour sediments using EDTA and citric acid aschelating agents,” J. Hazard. Mater., vol. 147, 2007, pp.
768-775.
[11] Sun B, Zhao F J, Lombi E, McGrath S P, “Leaching ofheavy metals from contaminated soils using EDTA,”
Environ. Pollut., vol. 113, 2001, pp. 111-120.
[12] Steele M C, Pichtel J, “Ex-situ remediation of ametal-contaminated superfund soil using selective
extractants,” J. Environ. Eng., vol. 124, 1998, pp.
639-645.
[13] Peter R W, “Chelant extraction of heavy metals fromcontaminated soils,” J. Hazard. Mater., vol. 66, 1999, pp.
151-210.
[14] Mn Q Y, Traina S J, Logan T J, “In situ lead
immobilization by apatite,” Environ. Sci Technol., vol.
27, 1993, pp. 1803-1810.
[15] Ma Q Y, Longan T J, Traina S J, “Lead immobilizationfrom aqueous solutions and contaminated soils using
phosphate rocks,” Environ. Sci. Technol., vol. 29, 1995,
pp. 1118-1126.
[16] Miretzky P, Fernandez-Cirelli A, “Phosphate for Pb
immobilization in soils: a review,” Environ. Chem. Lett.,
vol. 6, 2008, pp. 121-133.

741

742

Biogas Production from Tannery Sludge
A.Rajendra Prasad, A.Dhanalakshmi, S.Usha and A.Abdul Rahman
Sri Sai Ram Engineering College, Chennai, India.
Solid waste, emanating from the tannery
effluent treatment plant, is the primary sludge.
About 40% of this is organic matter. It is anappropriate source for conversion into biogas.
A substantial quantity of sludge from tannery
effluent treatment plant is currently disposed ineither safe landfills or dumped in the vacantlands, again causing a environmental hazard.

1.ABSTRACT
The present study was conducted using
Pilot plant scale Bio gas generator in CETP at
Chennai during 2012. Bio-gas was generatedusing two sintex tanks-anerobic digestion
chamber and floating gas holder respectively.
Cow-dung was used initially to generatebiogas,supplemented by tannery sludge. The
sludge generated from the Tannery effluent
treatment plant is pre-digested using
extremophiles and the slurry is used as a feed
to generate biogas.
Qualitative and quantitative analysis of
biogas were performed. The processed sludge
was fed into the biogas chamber to optimisetheemissionofbiogas.
Daily
production/emission of biogas is being
monitored. The Pilot plant now produces 700
litres of bio-gas and the continuous burning
time is 2 hours. The biogas is useful as a fuelsubstitute being non-polluting,renewable andCO 2 neutral.
There is no simple treatment to purify
contaminated or polluted air.It is therefore
necessary to reduce the pollutant load in the air
at the source itself.Methane which contains theother gases such as CO 2 ,NH 3 and H 2 S are now
treated and the efficiency of methane is also
increased.

3. OBJECTIVES:
The objectives of this initiative are,To evaluate the technology for generationof biogas from primary sludge from a Tannery
Effluent treatment plant under conditions
prevailing in the CETP at Pallavaram,
Chennai.
To evaluate the quality and volume of gas
produced with reference to raw material fed.
To obtain relevant data for assessing theconsumption of gas produced and to initiate
the treatment of biogas in order to increase themethane.
The overall solid waste generation fromTanneries is 59.3%. Out of this the fleshingconstitutes 13.9%.
3. MATERIALS AND METHODS:
Fabrication of Bio-gas pilot Plant requires
two Sintex tanks having 1000 litre capacitywith diameter of 110cm as fermenter and 750
litre capacitiy with diameter of 103cm as gas
holder. The top portion of the tanks have been
removed to accommodate the small tank of
750 litre capacity into big tank of 1000 litre
capacity.
Two holes in the bottom and one in the topof the fermenter tank have been made andappropriate PVC pipe lines have also been
provided. The bottom pipe serves to remove
the sludge in case of heavy accumulation ofsludge. Another hole is made with 90mm
capacity through which inlet pipe has been
provided. In top portion of the fermenter
another hole is provided with outlet to remove
excess amount of waste water.
The 750 litre capacity tank is provided withthe gas holder with a PVC pipe with holes
inside pipe for emission of gas through thepipe line from the gas holder and to the burner.

2. INTRODUCTION:
In tanneries, various types of solid wastes
are generated from the effluent treatment plantsuch as soak liquor, lime liquor and chrome
liquor. Among these effluents, soak liquor
forms the organic wastes predominantly
constituting the flesh, albumen, dung andparticularly the salts.
However, it is the lime liquor which
contributes as primary and secondary sludges
are being the major portion of the solid waste
need proper treatment and disposal.
However, in the case of chrome liquor,there
is recovery of chromium and being recycled
and hence it is not taken into consideration.
In the case of primary and secondary
sludges the disposal becomes major concern.
In addition to the sludge, there is considerable
amount of lime flesh which is generated assolid waste. The lime flesh after proper
washing of the lime in the flesh is used for thebiomethanisation.

742

- 743 3.1 METHODOLOGY

The tannery sludge is used effectively
to produce biogas in pilot plant scale. This
reduces the dumping of solid waste, minimize
solid waste generation, recycling of tannery
sludge. The biogas generation encourages to
produce biogas fuel utilizing huge quantity oftannery sludge waste effectively. The biogas
was collected, separated and purified.

During the initial stage of installation, thebiogas started accumulating in the gas holder.
Daily record of increase in the volume ofbiogas was also noted on the outer surface ofthe gas holder. When 75% of the gas holder israised, the gas was also tested. It is determinedthat 1000 litre fermentation tank, with 750 litre
gas holder will have 700 litre of biogas which
can be used as a fuel for 2hrs. The process is
now monitored. . The outlet gas showed blue
flame when external oxygen is mixed with it.
3.2 PREDIGESTIONSLUDGEOF

3.3 PRINCPLE OF BIOMETHANATION:
Anaerobic or methane formation isbrought about by extremophile -complex
bacteria.
The various forms of degradation ofcomplex organic matter in the various
anaerobic process go through three distinct
phases namely,
Hydrolysis and acidogenesis phase.
Acetogenesis phase.
Methanogensis phase.
The first phase is performed by hydrolysis
and results in a mixture of volatile fatty acids,
neutral compounds such as ethanol andgaseous products such as CO 2 and hydrogen.
The acetogenesis phase refers to acetateproduction and is brought about by bacteria,
which produce hydrogen. Thus this phase issensitive to the presence of hydrogen.
Methane – producing micro-organisms
bring about the methanogenesis aredenitrifiable. In the first, hydrogen and CO 2
form water and methane. In the second, acetate
is broken down into CO 2 and CH 4 .
It is the second method, which produces
about 70% CH 4 .
The composition of gas generated depends
on the composition of the substrate andconditions of operation. Gas production is thesimplest criterion to measure the quality ofdigestion and it depends on two main factors
namely temperature and retention time.

TANNERY

Tannery sludge was collected from the primary
settling tank and was used as the primarysource of energy for the biogas production.
However, during the study, sludge was alsocollected from the secondary and tertiary
settling tanks and the Carbon-Nitrogen ratio
was evaluated for selecting the primary sludge
for the purpose.
The same primary sludge was also converted
as compost in view of its nitrogen source.
Taking this into consideration the primarysludge was fermented using Bacterial
consortium. The bacterial consortium wasproduced from the culture of the cow dung as
the raw material and the molasses was used asa medium to culture the bacteria.
The bacterial consortium consists of various
types of anaerobes are used for the present
study to make the sludge into semi solid andslurry like organic biomass. This was fed into
the biogas chamber. Initially, 4 litres of slurry
was introduced into the chamber followed by
10 litres of chlorine-free water. However, afterten days of trial, the feed material wasincreased to 10 litres and consistent productionof biogas was obtained. Therefore, adding
slurry to the bio-gas plant was standardized.

- 743 -

- 744 -4. PROCESS FLOW:

Effluent fromprimarytreatment

Effluent fromsecondarytreatment

Wet sludge

Pre-digested sludgeRaweffluent

Impure biogas

Treated Bio-gas used asenergyFig.1- 744 -

Effluent fromtertiarytreatment

Extremophiles

- 745 -5. RESULTS AND DISCUSSION
Characteristics of sludge from CETP in Pallavaram, Chennai, Tamilnadu, India.
Sno.
Parameter*
CETP Pallavaram
1.
pH
7.7 -8.8
2.
Total Solids
45052.00
3.
Total Ash
27533.00
4.
Total volatile solids
17519.00
5.
COD total
22259.33
6.
COD soluble
2180.67
7.
Calcium as Ca
2000.00
8.
Sulphate as SO 4 2606.67
9.
Total Sulphide as S
370.67
10.
Total Phosphorus as P
43.29
11.
Ammoia Nitrogen
430.0012.
TKN
1314.67
13.
Sodium as Na
3953.33
14.
Total chromium as Cr
317.0015.
Chloride as Cl
9933.33*unit as mg/L other than pH
According to Thabaraj et al tannery waste
water could be treated to a level of
BOD30mg/litre through activated sludge process
at HRT 24 hrs. The MLSS concentration ofactivated sludge process was required to be at
4000 mg/litre for a removal of 83% COD and
92% Sulphide [(Cooper et al ) 1976]. The
primary clarification of tannery effluent withalum, ferrous sulphate or ferric chloride andsettling before aerobic biologic treatment

assisted in reducing COD by 93.6% (Cheda et al1984).
Thus the treatment of effluent generated sludge
and this sludge was used as feed for biogas plant
after making pre-digested sludge.

cost effective, without sacrificing the efficiency
of calorific value of methane gas.
However,using activiated carbon has produced useful
result but the efficiency is not encouraging.
Hence , it is also proposed to use different
molecular sieve and mesh size of activatedcarbon and to evaluate the suitability ofparticular molecular sieves and mesh size which
absorbs CO 2 under NTP having uniform
calorific value.
The traces of H 2 S needs removal. The
treated biogas dried using silica gel.
Alternatively Al 2 O 3 in case of low dew pointneed to be achieved.
An alternative method of drying using glycolor hygroscopic salt is also tried for absorption ofmoisture in the treated biogas. The result will be
evaluated for upscaling the technology in future.

5.1 TREATMENT OF BIOGAS
The impure biogas constituted 60% methane
rest being other gases. However the
moisturecontent increased in the treated biogas
after removal of other gases. Hence to remove
excess moisture activated carbon and glycol
was used. This has increased the calorific valueof methane.
The methodology for removal of CO 2 isusing common solvent water. The solubility of
CO 2 depends on pressure, temperature and pH.
In addition to it, Ca (OH) 2 was also used to form
CaCO 3 and CaS
However, it is proposed to use organic
solvents such as polyethylene glycol to make it- 745 -

- 746 -REFERENCES
1. Cheda, P.V., U.V. Mandelkar, B.K. Handa
and P. Khanna 1984. Joint wastewater
management for a cluster of tanneries at
Kanpur. 39th Industrial waste conference,
Purdue University proceeding pp 151-162

5.2 BIOGAS ANALTICAL DATAS.NO

CONTENT

1
2
3
4
5
6

CH 4
CO 2
NH 3
H2S
MOISTURE
OTHER
IMPURITIESIMPURE
% W/W
60
22.5
1.0
0.5
6.0
10

TREATED
% W/W
80
NIL
NIL
NIL
10
10

.

- 746 -

2. Cooper J.E., W.T. Happich, E.F. Melton and
J. Naghaki, 1976.
Activated Sludge treatment of tannery wastes.
JALCA. 71(1) 6-24.
3.Thabaraj, G.J., S.M. Bose and
Y. Nayudumma,1964. Utilisation of tannery
effluents for agricultural purposes. Indian
Journal of Environment
Health1:28–36.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Water Harvesting and Salinization Prevention by Capillary Barrier of Soil
1

Toshihiro Morii1, Mitsuhiro Inoue2, Kaoru Kobayashi3, Tetsuya Suzuki1and Takayuki Kawai1
Research Institute for Natural Hazards and Disaster Recovery, Niigata University, Niigata, Japan
2
Arid Land Research Center, Tottori University, Tottori, Japan
3
Research Institute of Technology, Tobishima Corporation, Chiba, Japan

ABSTRACT
A simple soil layer system which is composed of a finer soil layer underlain by a coarser soil layer provides acharacteristic property of capillary barrier. Water infiltrated into soil is suspended just above an interface between soil
layers due to a physical difference in unsaturated hydraulic properties of soil. As their roots can easily utilize the waterretained in the upper soil layer, agricultural plants grow even under water saving irrigation. Because the gravel layer
embedded in the soil cuts upward movement of water from groundwater, the capillary barrier of soil is expected toprovide an effective prevention of salinization which is caused by salt included in the groundwater. Two series of field
experiment are conducted on the sand soil where the gravel layer is embedded to investigate a practical effectiveness ofthe capillary barrier of soil in the water saving irrigation/ water harvesting and the salinization prevention.
Keywords: Capillary Barrier of Soil, Water Harvesting, Salinization Prevention, Field Experiment

1. INTRODUCTION
A simple soil layer system which is composed of a finer soil
(usually sand) layer underlain by a coarser soil (gravel) layer
provides a characteristic property of capillary barrier [1]-[4].
Water infiltrated into soil is suspended just above an interface
between soil layers due to a physical difference in unsaturated
hydraulic properties of soils as shown schematically in Fig. 1.
In the case of tilted interface, the suspended water flows
downward along the interface and a vertical movement ofwater into deeper soils below the interface will be prevented
within some length along the interface. This leads to apractical application of the capillary barrier system (CBS) to a
top cover of waste disposal landfill [5] and a slope repair [6].
In the case of a horizontal interface, plant can easily consume
the water retained in the upper soil layer for its growth [7].
Because of this, the CBS may be well employed for an
Rainfall/ Irrigation

Sand

Gravel
Any soil

Moisture content

Depth

0

Water infiltrated andaccumulated above
the interface between
sand and gravel

Fig. 1 Schematic diagram to show water accumulation
above an interface between finer soil (sand) and coarser
soil (gravel) caused by capillary barrier of soil.

effective water harvesting and saving irrigation in an area
where water resource is scarce. As the gravel layer of the CBS
embedded in the soil cuts upward movement of water fromgroundwater, it may be expected that the CBS will provide aneffective prevention of salinization which is caused by salt
included in the groundwater.
In the study, two series of field experiment were conducted on
the CBS soil where the gravel layer was embedded in the sandsoil in order to investigate a practical effectiveness of the
CBS in water harvesting and salinization prevention. In thefield experiment for water harvesting [7], experiment plots of
the CBS soil were constructed, green vegetables were plantedon them, and soil moisture contents were measured after
seeding to harvesting of the plant. The water retained in theupper sand layer was observed, and its effect on the plant
growth was quantified by a statistical analysis of the plant
harvest. In the series of field experiment for salinization
prevention, a number of the CBS soil were constructed inpolyester containers, and connected to a water tank which
could supply an artificial groundwater, 10,000ppm of NaCl,to the CBS soils with a constant head of water. Soil moisture
content and electric conductivity in the upper sand layer, thegravel layer and the lowest sand layer overlain by the gravel
layer were measured after plant to harvest of the green
vegetable. It was observed that the gravel layer embedded in
the CBS soil at some depth above the groundwater surface
can well prevent the water from moving upward from thegroundwater surface. This excellent interception of upward
water movement was also observed by the plant death due tothe salinization in the soil without the gravel layer.
Both series of field experiment reveals that the CBS is wellapplicable to the water harvesting and the salinization
prevention, which may suggest some new progress ofgeotechnical engineering to green technology.

747

2.1 Experiment Plots
Six experiment plots, each 70 cm wide and 90 cm long, wereconstructed to compare plant growth in the greenhouse as
shown in Fig. 2. The sand soil was excavated into a depth ofabout 40cm, and veneer boards were placed to retain soil walland to prevent lateral flow of soil water during theexperiment. Then the sand soil was filled back into the firsttwo of six experiment plots, S-1 and S-2 in Fig. 2, andcompacted uniformly along all the depth. In other two
experiment plots, G10-1 and G10-2, after the sand soil was
filled back and compacted along the depth 40 to 20cm, agravel layer 10 cm in thickness was placed and compacted,
and then the sand soil was filled back and compacted
uniformly from the depth of 10cm to the soil surface. In theremaining two experiment plots, G20-1 and G20-2, a gravel
layer 10cm in thickness was placed and compacted between
the depth of 30cm and 20cm. A polyester nonwoven sheet
was placed over the compacted gravel layer so that the sandsoil particles did not fall into voids formed in the gravel layer.
This sheet was tested to be completely permeable. Numerals
1 and 2 given after a hyphen in the experiment plot number
mean a repetition for experimental design.
Seed lines

Plane

700

Veneer boards
700

S-1

G10-1

G20-1

S-2

G10-2

G20-2

900

900

700

＠100

400

Access tubes for moisture measurement
Sand
SectionGround surface
Sand
Veneer boards

GravelNote: Length and depth are in mm.

Fig. 2 Plane and sectional configuration of experimentplots constructed on the sand soil.
2.2 Materials
Fig. 3 shows grain size distribution curves of sand andgravel employed in the experiment plots. The sand has
less-5% fine and coarse fractions; the gravel, commerciallyavailable, is siliceous with a mean particle size of 5 to 6mm.
Soil particle density, dry density of compacted soils andPercent finer by mass, %2. FIELD EXPERIMENT FOR WATER HARVESTING

100
50
0
0.01

Sand

Gravel0.1
1
10Soil particle diameter, mm

100

Fig. 3 Grain size distributions of sand and gravel used toconstruct the experiment plots.
Table I Soil particle density, dry density of compacted soils
and hydraulic conductivity.
Properties

Sand

GravelSoil particle density, Mg/m3
2.626
2.653
Dry density of soil, Mg/m3
1.46-1.56
1.69-1.80
Hydraulic conductivity*, m/s
0.99×10-4
1.04×10-2*Corrected to water temperature at 15 degrees Celsius.

hydraulic conductivity of sand are summarized in Table I.
Dry density of soil was calculated by dividing a mass of soil
compacted into the layer by a volume of the layer. Saturated
hydraulic conductivity of sand was determined byinterpolating the results of laboratory seepage test to the meanvalue of dry density of sand layer, that is, 1.51Mg/m3.
Saturated hydraulic conductivity of gravel was measured by
the laboratory seepage test under unit hydraulic gradient.
Relationships of negative pore pressure with volumetricmoisture content of sand and gravel were measured by alaboratory soil column test, and are plotted in Fig. 4(a).
Soil-water characteristic curves are determined by using van
Genuchten equation [8]. Fig. 4(b) shows unsaturated
hydraulic conductivity of sand and gravel estimated from Fig.
4(a) using the van Genuchten equation. It is found that, when
the negative pressure head just above the interface between
sand and gravel decreases to 3 to 4cm due to accumulation ofwater, the hydraulic conductivity of gravel becomes larger
than that of sand and the accumulated water will begin to
infiltrate into the gravel layer. But, because nearly 100%
saturation within the sand layer as estimated from Fig. 4(a)rarely occurs under evaporation condition with a restrictedsupply of water, the breakthrough of accumulated water into
the lower gravel layer may not either.
2.3 Planting
Potherb Mustard was seeded at 5cm intervals along seed lines
shown in Fig. 2(a) on 11 July, thinned on 28 July in all theexperiment plots, and harvested on 14 August to measureheight, fresh mass and dry mass of plant. Water
corresponding to 100 % of pan evaporation measured the day
before was supplied by a watering can every morning during
a period from seeding to thinning, and 50 % during thinningto harvesting. Volumetric moisture content of soil in theexperiment plots were measured by using Profile Probe748

Seeding on 11 July(a) SWCC

Evaporation and watersupply, mm/day

0.4
Sand (D)Sand (W)

0.2
0.0
1

Gravel (W)

1
10-1

10

Negative pressure head h, cm

Sand (W)
Gravel (W)

-4

1

10

Negative pressure head h, cm

100

Fig. 4 Soil-water characteristic curves (SWCC) of sandand gravel measured by laboratory soil column test andestimated by using van Genuchten equation in (a).
Unsaturated hydraulic conductivities of sand and gravel
are estimated in (b).
PR2/4 with a moisture meter HH2 (Delta-T Devices Ltd.), at
positions given by painted circles in Fig. 2 along the depth ofsoil surface to 40 cm at 10 cm intervals. Relationships
between output of PR2/4 and volumetric moisture content ofsoil were calibrated along the probe in the laboratory.
2.4 Results
Evaporation was measured to be about 1 mm to 6 mm during
the field experiment as shown in Fig. 5. The volumetricmoisture contents were measured along the depth of soil atweekly intervals after seeding, and given in Fig. 6 together

10

1 42
5

3

0

Sand

20
30
(a) S-1

40
0.0
0.1
0.2
Volumetric moisture content
θ, cm3/cm3

Depth z, cm

Depth z, cm

0

1 : 12 Jul
2 : 19 Jul
3 : 26 Jul
4 : 2 Aug5 : 9 Aug

1

10

Water supply

Evaporation

5
0

7
14
21
28
Days after seeding on 11 July35

Fig. 5 Evaporation measured during the experiment, andwater supply.

Sand (D)

10-3

50% supply

100% supply

0(b) Unsaturated hydraulicconductivity

10-2

10

100

Harvesting on 14 Aug.
Thinning on 28 July

15

with the position of the gravel layer. In Fig. 6(a) which shows
soil water profiles in the experiment plot S-1 composed ofsand alone, it is seen that the water supplied on the soil
surface infiltrates into the deeper position. On the contrary, inFig. 6(b) and Fig. 6(c) in the experiment plots with the gravel
layer at depth 10cm and 20cm respectively, it is found thatwater supplied on the soil surface infiltrates into the sandlayer and then accumulates just above the interface between
sand and gravel layers with less infiltration into the gravel
layer. It may make sense to think that the plants can easily
absorb the accumulated water to consume for their growth.
2.5 Effect of Accumulated Water on Plant Growth
An effect of water accumulated due to the CBS on the plant
growth was investigated by using a statistical analysis. In Fig.
7, mean values of plant height excluding root length, freshmass and dry mass of plant are plotted in thick bar together
with standard deviation given by thin line. The number ofplant harvested and analyzed is 44, 43 and 51 in theexperiment plot S (that is S-1 plus S-2), G10 (G10-1 plus
G10-2) and G20 (G20-1 plus G20-2), respectively. Alphabets

5 4 2 30Sand

10Gravel20

Depth z, cm

Volumetric moisture
contentθ, cm3/cm3
Unsaturated hydraulicconductivity K, cm/s

Note that D: drainage and W: wetting.

0.6

1

5

24

3

10

Sand

20

Sand

30
(b) G10-1

40
0.0
0.1
0.2
Volumetric moisture content
θ, cm3/cm3

Gravel30(c) G20-1

40
0.0
0.1
0.2
Volumetric moisture content
θ, cm3/cm3

Fig. 6 Volumetric moisture content along depth of the experiment plots measured during a period from seeding toharvesting. Data are plotted at weekly intervals.
749

Sand

30

20
10
0

b
a

c

S G10 G20
Experiment plot(a) Plant height4
Fresh mass, g

30Fresh mass, g

Plant height, cm

40

20

10

b
a

0

b

3
2
1
0

S G10 G20
Experiment plot(b) Fresh mass of plantbba
S G10 G20
Experiment plot(c) Dry mass of root

Fig. 7 Multiple comparisons to show the effect of the CBS on the plant growth. The experiment plots de-noted by different
alphabets such as “a” and “c” are statistically different each other with a significant level of 5%.
‘a’, ‘b’ and ‘c’ given in Fig. 8 show the result determined by
the multiple comparison analysis based on Scheffe's method.
The experiment plots denoted by different alphabets such as‘a’ and ‘c’ are statistically different each other with asignificant level of 5%. According to this analysis, it is foundin Fig. 7 that the plant height, the fresh mass and the dry massof plant grown in the experiment plots G10 and G20 are
significantly larger than those in the experiment plot S. Thissuggests water accumulated in the sand layer due to thecapillary barrier of soil was consumed effectively by the plantfor its growth. It is also found that the plant height in theexperiment plot G10 is significantly higher than that in G20
in Fig. 7(a), and the fresh mass of plant in G10 are larger thanthose in G20 in Fig. 7(b). This is because, as the plant roots
were observed to grow thick within the soil about 10cm deep,
they could not effectively absorb water accumulated within
the sand layer in the experiment plot G20.
3 FIELD EXPERIMENT FOR SALINIZATIONPREVENTION
3.1 Experiment Plots
In the field experiment to investigate the salinization
prevention effect of the CBS, a polypropylene container box
61 cm long, 41 cm wide and 31.5 cm deep was used toconstruct the experiment plot as shown in Fig. 8. In the plot
G25 and G50, gravel layer 2.5 cm and 5 cm in thickness wereembedded in the container boxes, respectively. No gravel
layer was placed in the container boxes in the plot S. A line
linking three container boxes is connected to a water supply
tank which can supply groundwater into the container boxes
with constant head of water.
The sand soil was excavated into the depth of about 30 cm,
and the container boxes were placed. Then the sand soil was
filled back to the container box and compacted uniformly into
the thickness of 16.5 cm, that is, along the depth 31.5 cm (thebottom of the container box) to 15 cm. In three container
boxes along the plot G25, gravel layer 2.5 cm in thickness
750

Plane

Reservoir

Supply tankwith constant head

G50(1)

G50(2)
2
1
3

S(1)G25(1)

S(2)

G25(2)
8
7
9

5
4
6

Sensors 5TE

G50(3)

G25(3)

Gravel layer

Section
3
1

S(3)

2

Thickness ofgravel layer= 50mm

6
4

5

9
7

8

Thickness ofgravel layer= 25mmFig. 8 Experiment plots to investigate the salinization
prevention by the CBS. See also Fig. 10.

0.2

3

0.1

Electric conductivity, ds/m

0.0
0
4
3

2
10 20 30 40
Time after set-up, days
G50(2)

50

1

2
1
0
0

32
10 20 30 40
Time after set-up, days

50

0.5

4

Volumetric moisture
content, cm3/cm3

0.3

0.4

S(2)

5

0.3

6

0.2
0.1
0.0
0
4
3

10 20 30 40
Time after set-up, days
S(2)

4

2

5

1
0
0

50

6
10 20 30 40
Time after set-up, days

50

0.4

G25(2)

7

0.3
0.2

9

0.1
0.0
0

Electric conductivity, ds/m

0.5
1Electric conductivity, ds/m

0.4

G50(2)

Volumetric moisture
content, cm3/cm3

Volumetric moisture
content, cm3/cm3

0.5

4
3
2

8

10 20 30 40
Time after set-up, days50

G25(2)
7

1
0
0

98

10 20 30 40
Time after set-up, days

50

Fig. 9 Volumetric moisture content (upper three graphs) and electrical conductivity (lower three graphs) measured by the
5TE sensors in the capillary barrier soil G25 and G50 and in the sand soil S. Numerals 1 to 9 indicate the number of the 5TE
sensor embedded in the soil as shown in Fig. 8.
was placed and lightly compacted, the same polyester
nonwoven sheet as one used in the preceding experiment was
spread on the gravel layer, and the sand soil was again filledback and compacted uniformly into the thickness of 10 cm. In
three container boxes along the plot G50, the gravel layer 5
cm thick, the polyester nonwoven sheet and the sand 10 cm
thick were placed in sequence. In the plot S, the sand soil was
continuously filled back and compacted uniformly from theremaining depth 15 cm to the soil surface. The sand andgravel are the same as ones used in the field experimentdescribed in Section 2, and their grain size distribution andthe material properties are given in Fig. 3 and Table I.
The groundwater level was kept 11.5 cm above the bottom of
the container boxes, with the result that vertical distance
between the groundwater level and the lower interface of thegravel layer was 5 cm. As a water-entry pressure head of thesand soil is estimated to be about 10 cm along the wetting
curve, this value of 5 cm is sufficiently small to investigate
the effect of the gravel layer on the upward movement ofwater from the groundwater. Sodium chloride was used tosupply saline groundwater, 10,000 ppm in mass.
3.2 Planting and Measurements
The groundwater was supplied to all the container boxes on 3
September. The young vegetables of Potherb Mustard wereplanted into the container box, and harvested on 21 October.
750 to 1,000 cm3 of water per a container box was supplied
every morning by using the watering can. During theexperiment, the volumetric moisture content and the electricalconductivity were monitored by using the water content, EC
and temperature sensor 5TE manufactured by Decagon
Devices, Inc. As shown in the section of Fig. 8, the 5TE
sensors No.1, 4 and 7 were embedded just below the lower

interface of the gravel layer or at the position of 16 cm above
the bottom of the container box; No. 2, 5 and 8 just above thelower interface of the gravel layer or at the position of 17 cm
above the bottom of the box; No. 3and 6 just above the upperinterface of the gravel layer or at the position of 22 cm above
the bottom of the box and No.9 at the position of 19.5 cm
above the bottom of the box.
3.3 Results
The volumetric moisture content and the electricalconductivity measured by the 5TE sensors in the capillarybarrier soil, the plots G25 and G50, and in the sand soil, theplot S, are given in Fig. 9.
At around 1 week after the saline water was supplied into theexperiment plots with the constant level of groundwater, boththe volumetric moisture content and the electricalconductivity measured in the sand soil below the gravel layer
increase rapidly in the experiment plots G25 and G50, and ataround 2 weeks the volumetric moisture content reached its
saturated value (No. 7 in the G25 and No. 1 in the G50).
However, during this 2 weeks and from then, no or negligible
change in the volumetric moisture content and the electricalconductivity is found in the gravel layer as given by No. 8 in
the G25 and by No. 2 in the G50, which means no upward
movement from the groundwater into the gravel layer. Small
increases in volumetric moisture content measured by No. 9
in G25 and by No. 3 in the G50 are caused by continuous
accumulation of infiltrated water just above the interface
between the upper sand layer and the gravel layer as shown in
Fig. 6. The electrical conductivity which has a strong
relationship with pore water salinity indicates almost zero justabove the interface of the CBS in the G25 (No. 9) and G50(No. 3) during the experiment period. It may be concluded

751

well prevent the water from moving upward from thegroundwater surface. This excellent interception of upward
water movement was also observed by the plant death due tothe salinization in the soil without the gravel layer.
Both series of field experiment reveals that the CBS is wellapplicable to the water harvesting and the salinization
prevention, which may suggest some new progress ofgeotechnical engineering to green technology [9].
5 ACKNOWLEDGMENTS
The present study is supported by the Grant-in-Aid for
Scientific Research (B), No. 22380127, made by the Ministryof Education, Science, Sports and Culture of Japan, and by
the Joint Research Grant made by the Arid Land ResearchCenter, Tottori University, Japan. Misses M. Sugisaki and K.
Takahashi and Mr. S. Furukawa are gratefully acknowledged
for their help in conducting the field experiment.

Fig. 10 Vegetables at the end of the experiment. Although
the plants grew successfully in the G25 (the right row of
the container boxes) and in the G50 (the left row), they
withered completely in the S (the central row) due to thesoil salinization.
from these measurements shown in Fig. 9 that the saline water
was completely prevented from moving upward from thegroundwater into the gravel layer and into the upper sandlayer where the vegetables were planted. In the experimentplot S without the gravel layer, both the volumetric moisture
content and the electrical conductivity increase according to
the upward movement of saline water from the groundwater
because there was no prevention of water movement by thecapillary barrier.
Fig. 10 shows a picture of the plots taken at the end ofexperiment. All the vegetables planted in the experiment plot
S without the gravel layer completely withered due to the soil
salinization. Be sure that, because the soil was not fertilized to
avoid an effect on measurement of the electrical conductivity
by the 5TE sensors, plant growth was not sufficient in theG25 and the G50 in Fig. 10.
4 CONCLUSIONS
Two series of the field experiment were conducted on thecapillary barrier soil in order to investigate a practicaleffectiveness of the CBS in water harvesting and salinization
prevention. In the field experiment for water harvesting,experiment plots of the capillary barrier soil wereconstructed, green vegetables were planted on them, and soil
moisture contents were measured after seeding to harvestingof the plant. The water retained in the upper sand layer wasobserved, and its effect on the plant growth was quantified bya statistical analysis of the plant harvested. In the series offield experiment for salinization prevention, a number of thecapillary barrier soil were constructed in polyester container
boxes, and connected to the water tank which could supply an
artificial groundwater, 10,000ppm of NaCl, to the experimentplots with a constant head of water. Soil moisture content andelectric conductivity in the upper sand layer, the gravel layer
and the lowest sand soil overlain by the gravel layer were
measured after plant to harvest of the green vegetable. It wasobserved that the gravel layer embedded in the capillarybarrier soil at some depth above the groundwater surface can

6 REFERENCES[1] Ross B, “The diversion capacity of capillary barriers,”
Water Resources Research, vol. 26, No. 10, 1990, pp.
2625-2629.
[2] Walter M, Kim J, Steenhuis T, Parlange J, Heilig A,
Braddock R, Selker J and Boll J, “Funneled flowmechanisms in a sloping layered soil: Laboratory
investigation,” Water Resources Research, vol. 36, No.
4, 2000, pp. 841-849.
[3] Rahardjo H, Krisdani H and Leong E, “Application of
unsaturated soil mechanics in capillary barrier system,”
Proceedings of the Third Asian Conference on
Unsaturated Soils, 2007, pp. 127-137.
[4] Morii T, Takeshita Y, Inoue M and Matsumoto S,“Alternative measures for soil slope stability using
capillary barrier of soil,” Proceedings of the FourthAsia-Pacific Conference on Unsaturated Soils, 2009, pp.
319-324.
[5] Morii T, Kobayashi K, Matsumoto K, Suzuki T and
Kawai T, “Practical application of capillary barrier ofsoil into a shallow land waste repository,” Proceedings ofthe Fifth China-Japan Geotechnical Symposium, 2012.
(to be published)[6] Rahardjo H, Satyanaga A and Leong E, “Unsaturated soil
mechanics for slope stabilization,” Proceedings of theFifth Asia-Pacific Conference on Unsaturated Soils, vol.
1, 2011, pp. 103-117.
[7] Morii T, Inoue M, Komatsu G, Kadoguchi R and
Takeshita Y, “Effective water harvesting using capillarybarrier of unsaturated soils,” Proceedings of the Fifth
Asia-Pacific Conference on Unsaturated Soils, vol. 2,
2011, pp. 857-860.
[8] Stephens D, Vadose Zone Hydrology. Florida: CRC
Press, Inc., 1996, ch. 4.
[9] Morii T, Inoue M and Takeshita Y, “Green technologicaleffect of capillary barrier of soil on agricultural plant
growth,” Proceedings of the Second Japan-Korea Joint
Workshop on Unsaturated Soils and Ground, 2011, pp.
249-256.

752

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Study of the Performance of anaerobic digestion of greasy skim as a renewableenergy source
Salam J. Bash Al-Maliky
AlMustansiriya University, Iraq

ABSTRACT
Five laboratory scale batch reactors (each of 10 L working volume) were used to test the effect of different Induced
Air Flotation IAF skim to working volume ratios; namely 0.5, 1.5, 2.5, 4.0, 5.0:10 (R1-R5), with the rest of working
volume was filled with livestock manure (as inoculums). These reactors were operated at two different temperatures(25C and 45C), to determine the temperature effects on digestion performance. Chemical Oxygen Demand (COD)
removal efficiencies of 72.6%, 68.6%, 60.1%, 52.1%, and 43.25% were achieved for R1-R5 respectively, attemperature of 25C. These removal efficiencies were significantly improved at temperature of 45C, where efficiencies
of 91.2%, 81.5%, 72.1%, 60.7% and about 50% were achieved for these reactors. Very promising biogas production
rates of 4.5- 5.8 l/day were determined at operation temperatures of 45C (compared with 3- 4 l/day at 25C) for the
majority reactor life times.
Key words: Anaerobic digestion, biogas, digestion temperature, Induced Air Flotation, substrate composition.
1. Introduction
Many recent researches and studies had demonstrated thegreat feasibility of applying the IAF approach for theremoval of Oil and Grease (OG) from the effluents ofmunicipal and industrial activities (Al-Maliky, 2010;
Arnold and Stewart, 2008; Pan Li and Hideki Tsuge,2006). IAF may simply be explained as a treatment
process often used to remove this insoluble particulatematter from the wastewater, essentially to reduce the OG
constituent, Biological Oxygen Demand (BOD), COD
and Suspended Solids SS loads of the wastewater toavoid surcharges when discharging to surface water ormunicipal wastewater treatment plants. The process isachieved by injecting air bubbles into the water orwastewater in a flotation tank or basin. The small bubbles
adhere to the suspended matter causing the suspended
matter to float to the surface of the water where it may
then be removed by a skimming device. The feed water
to the IAF float tank is often dosed with a coagulant toflocculate the suspended matter (Hayatdavoudi, 2006).
Many studies had suggested various design and operation
parameters for the IAF unit to achieve high performancesin oil/water separation, reduction of BOD, TSS and COD(Al-Maliky, 2010; Van Ham, Behie and Svrcek, 2009),
and in turn the production of skim sludge at the upper
layer of a typical IAF unit that is characterized by highconcentrations of OG, COD and TSS. Although thatskim is typically rendered or land applied, but its poor
natural degradability and high energy potential contents
make it a potential candidate for anaerobic digestion.
This approach needs careful handling to overcome thetendency of high OG concentrations to form a scum layeron the digester’s liquid surface due to the poor solubilityof OG at typical digester temperatures and the tendency753

to form a separate phase above the aqueous digester
contents (Halalsheh et al., 2005). Suggestions to handle
the later obstacle were either to reduce the IAF skim
amounts to be digested or the use of some inoculums
such as the agricultural wastes that improve microbial
nutrition and provide pH buffering capacity (Bouallagui
et al., 2009) .
Anaerobic treatment consists of the decomposition oforganic material in the absence of free oxygen and thisprocess produces biogas enriched in methane, carbon
dioxide, ammonia and traces of other gases and Volatile
Fatty Acids (VFA) within the reactor. The anaerobic
treatment process has been employed in several
developed countries with the aim of bio-stabilizing
fermentable organic waste produced by rural and urban
activities (Fernández, Pérez and Romero, 2010).
The main coal of this paper was to study the feasibility ofan anaerobic digestion process as a final treatment of theskim of IAF unit that handles dairy industry`s waste
water, and the potential of such process as renewableenergy source, under different set ups of digest mixtures(IAF skim and livestock manure) and operation
temperatures. Livestock manure was collected fromGreen Grass Farms, Athens, Ohio, USA.
1.1 Characteristics of IAF skim
The characteristic of having high COD concentration(425 g/L) as shown in Table 1 gave the IAF skim a highpotential of biogas production and encouraged increasing
the ratio of IAF skim to manure in the reactor feed tosupply a promising source of economic renewableenergy, subject to the constraints due to microbial
nutrition and pH buffering.
1.2 Experimental Setup
Five batch reactors (10 L working volume) were adopted
for the test of anaerobic digestion parameters, asillustrated in Fig.1. Each reactor has an independent
rotary blade mixer LIGHTNIN SPX so as to maintain
uniform moisture content and to homogenize the soluble
substrate and bacterial distribution. Also, an internal
DUREX coil heater was inserted in each reactor, withthermostat so as to supply the required heat to meet thetest`s operational temperature.
Under Each set of operational parameters, reactors were
fed initially with 10 L of diluted sludge (animal manure
diluted with fresh water) until the appearance of somelow biogas production rates for successive operation
days and in turn the obtained digested sludge would
serve as inoculums for reactors.
Sludge masses were made of dewatered IAF unit skim
and kept at -20 C prior to use as the feed for reactors.
Initially, the pH of digested sludge was adjusted to therange 6- 8 via 2N solution of Sodium Hydroxide. ThispH was monitored continuously to make sure it is in that
range.
Analysis of COD, TSS and pH were determinedaccording to the standard methods (APHA, 2010). The
biogas production was determined by the displacementof water in a graduated column.
Digest substrates were made of 0.5, 1.5, 2.5, 4.0 and
5.0:10, IAF skim to the total working volume (each in
one reactor; R1, R2, R3, R4 and R5 respectively) at eachof the test temperatures (25C and 45C).These
combinations were initially mixed via a commercial
laboratory blender LBC15.
The five reactors were operated for a period of 30 days
under each certain set of operational parameters.

OG content of the IAF skim; e.g., at higher temperatures,
OG tends to be more soluble that in turn enhances its
biodegradability.
Regarding the TSS constituent, the performance trend of
reduction was almost similar to that of COD, which is areference to the degradable organic composition of the
majority of these solids. At operation temperature of
25C, the reduction percentages were 83- 49% for thereactors R1-R5, respectively. The same scenario of high
COD removal enhancement at operation temperature of
45C, was repeated with the TSS, where these percentages
had the values of 95- 77% for the same mixing ratios.

2. Results and Discussions
Although disturbances in reactors performance were
expected for the early periods of operation, pH values
were mainly within the acceptable rang of 6-8 as shownin Fig.2 for the majority of reactors life times, whichindicates well buffering balance of substrate mixtures.

2.2 Biogas Production
Biogas production in Fig.4 (a) and (b) has shown
significant increase via operating the different reactors
with the mixed substrates compared to that during theanaerobic digestion of the diluted livestock manure only,
as shown in Fig.5. Also, these results had clearly
demonstrated, that higher IAF skim (higher COD) has
the tendency of higher biogas production at both
operation temperatures (25C and 45C). The greater
biodegradability and energy potential of carbohydrates
that constitute IAF skims might be the main causal for
that, in agreement with Thomas et al. (2007).
Highest biogas production rate was measured for R2(about 4.2 l/day), in comparison with that for R1 (about
3.7 l/day), that hosted the highest IAF skim: working
volume ratio at an operation temperature of 25C (Fig.4(a)). This may be attributed to the higher OG content andsalt concentration in the later mixture that inhibits
biological transformation at that temperature, and hence,
adversely affects the biological activity. That decrease of
biogas production rate at higher concentration of IAF
skim was overcame by operating the digestion reactors at
45C, (rates of 5.8 l/day and about 5.0 l/day were achieved
with R1 and R2 respectively), due to the enhanced
solubility at high operation temperatures as shown in
Fig.4 (b).

2.1 Digestion Performance
In general, it takes a period of time for the reactor to
acclimate and start being effective in the sense of COD
removal due to long-chain fatty acids released from thehydrolysis of triglycerides constituents of OG, and thelength of this lag period depends upon the IAF skim:working volume ratio; the higher it was, the longer thelag period. From another side, Fig. 3 (a) indicates thathigher IAF skim portions had led to lower COD removal
efficiency; that is, while the maximum achieved
efficiency with reactor R1 was about 72.6%, it did notexceed 43.25% with reactor R5, when operated attemperature of 25C (68.6%, 60.1% and 52.1% for
R2-R4 respectively). Remarkable enhancements wereachieved for the COD removal efficiencies that reached
the values of 91.2%, 81.5%, 72.1%, 60.7% and about
50% for R1- R5 respectively, by operating the reactors at
higher temperature of 45 C, as shown by Fig.3 (b). Thismay be attributed to the thermophilic characteristic of

3. Conclusions
The anaerobic digestion of wastes that are characterizedby high constituency of COD and OG (IAF skim for thisstudy) was tested via co-digestion with livestock manure
at different mixing ratios and two operation temperaturesin laboratory scaled batch reactors. Good COD removal
efficiencies (about 72.6%- 68.6%) were achieved while
using low portion of these wastes (5% - 15% of the total
working volume, respectively) at operation temperatureof 25C. This removal efficiency was significantly
improved to the values of 91.2%- 81.5% for the same
mixing ratios, at operation temperature of 45C, whichleads to the recommendation of operating such digesters
at high temperatures to make benefit of the thermophiliccharacteristic of such wastes. That later conclusion was
so demonstrated by the great enhancement of the biogas
production rate of such reactors when operated at hightemperatures; while the highest biogas production rate754

was in the range of 3-4 l/day at temperature 25C, it was
raised to the range of 4.5- 6 l/day at the temperature 45C.
4. References
Al-Maliky Salam B., (2010). Effect of geometricaldimensions and waste water temperature on theperformance of an induced air flotation unit for thetreatment of industrial waste water. Journal of Modern
Applied Science Vol.4 No.6, pp. 14-19.
APHA (2010), Standard Methods for the Examination of
Water and Wastewater, Edited by Lenore S. Clesceri,
Arnold E. Greenberg and Andrew D. Eaton. (20thEdition).
Arnold, K.E., Stewart,M., (2008). Surface production
operations-design of oil handling systems and facilities.
(Vol. 1, 3rd ed.), Gulf Publishing Co, Houston, Texas.
Bouallagui, H., Lahdheb, H., Ben Romdan, E., Rachdi,
B., Hamdi, M., (2009). Improvement of fruit andvegetable waste anaerobic digestion performance andstability with co-substrates addition. Journal ofEnvironmental Management 90, 1844–1884.
Fernández J., M. Pérez b, Romero L.I., (2010), Kinetics
of mesophilic anaerobic digestion of the organic fractionof municipal solid waste: Influence of initial total solid
concentration. Journal of Bioresource Technology, 101,
6322–6328.
Halalsheh, M., Koppes, J., den Elzen, J., Zeeman, G.,
Fayyad, M., Lettinga, G., (2005). Effect of SRT andtemperature on biological conversions and the related
scum forming potential. Journal of Water Resources. 39,
2475–2482.
Hayatdavoudi A., (2006). Removing Oil and Grease
from Produced Water Using Micro-Bubble Flotation
Technique. Offshore Technology Conference, Texas,
USA.
Pan Li and Hideki Tsuge (2006). Water Treatment by
Induced Air Flotation Using Microbubbles. Journal ofchemical engineering of Japan, Vol.39, No.8, pp.
896-903.
Thomas, A., Barbara, A., Vitaliy, K., Werner, Z., Karl,
M., Leonhard, G., (2007). Biogas production from maize
and dairy cattle manure: Influence of biomass
composition on the methane yield. Agriculture
Ecosystems and Environment 118, 173–182.
Van Ham Niel, Behie Leo A. , Svrcek William Y. (2009).
Effect of air distribution on the induced air flotation offine oil in water emulsions. The Canadian Journal ofChemical Engineering, Vol. 61 Issue 4, pp. 541 – 547.

755

Table 1 Characteristics of IAF skim under study
CharacteristicpH
COD (g/l)
TSS (g/l)
OG (mg/l)

Amount4.5- 5
425
120
870

5.00

◊◊◊ R1+++ R2∆∆∆ R3∆∆∆ R4○○○ R5

80.00

60.00

Biogas production rate l/day

COD Removal efficiency %100.00

40.00

20.00

0.00
10.00

20.00

(a)

3.00

2.00

1.00

30.00

0.00

Day

10.00

a)

100.00

20.00

30.00

Day
6.00

80.00

Biogas production rate l/day

COD Removal efficiency %

4.00

0.00
0.00

60.00

40.00

20.00

0.00

4.00

2.00

0.00
0.00

(b)

◊◊◊ R1+++ R2∆∆∆ R3∆∆∆ R4○○○ R5

10.00

20.00

30.00

Day

b)

Figure 3. Daily COD removal efficiencies for the reactors
R1-R5at; (a) Temperature of 25C. (b) Temperature of 45C.

0.00

10.00

Day

20.00

30.00

Figure 4. Daily biogas production rate for the reactors
R1-R5at; (a) Temperature of 25C. (b) Temperature of 45C.

Biogas production rate l/day

5.00

4.00

3.00

2.00

1.00

0.00
0.00

10.00

Day

20.00

30.00

Figure 5.Biogas production rate for anaerobic digestion of
livestock manure only.
756

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Seismic evaluation of NARGES soil nailed wall under Cyclic loading and
Pseudo Static forces

1.
2.
3.

Ardalan Akbari Hamed1 , Mohammad Moghadaripour 2, Ali Ghozat3
Islamic Azad University, Central Tehran Branch, Tehran, I.R .Iran
International Institute of Seismology and Earthquake Engineering, Tehran, I.R. IranShiraz University, Shiraz, I.R .Iran

Abstract
The more widespread use of soil nailed walls contemporary is due in large part to the demand for theconstruction of high-rise structures and shopping malls. Design of retaining wall needs the complete knowledgeof earth pressures for both active and passive conditions. Moreover under earthquake condition, the design
requires special attention. In this paper NARGES soil nailed model has been employed to study the seismic
behavior. It is attempted to achieve a relation between ground seismic coefficient and ground motion amplitude
to conclude the value of seismic coefficient and seismic study of this wall. Two numerical analysis methods (i.e.
cyclic time history and pseudo-static) are applied to simulate the seismic behavior. The computational results
show that seismic coefficient is in form of a limited range of ground seismic normalized horizontal acceleration.
Keywords: soil nailed walls, natural frequency, resonance, seismic coefficient

1.Introduction

The objectives of current paper are to observe thestability of a specific wall under seismic loads,
determine the value of seismic coefficient andseismic study by pseudo static method, it is
attempted to use the gained relation between
ground seismic coefficient and ground motion
amplitude.

Soil nailing consists of the passive reinforcement(i.e., no post-tensioning) of existing ground by
installing closely spaced steel bars (i.e., nails),
which are subsequently encased in grout. As
construction proceeds from the top to bottom,
shotcrete or concrete is also applied on the
excavation face to provide continuity [1]. Deep
excavations and retaining structures are constructedin the city of Tehran at different locations of thecity due to the recent demand for the constructionof high-rise structures having various basements.
Tehran is located at a very seismically active
region and a major earthquake magnitude ofMw>7.0 is expected to occur. It is well known thatflexible earth retaining structures in cuts, such as
soil nailed walls, offer a great advantage forseismic conditions. As a result many soil nailed
walls having of different heights have been
constructed recently in the city.
The pseudo-static method is routinely used for theseismic stability analysis of soil nail walls. In thismethod, an equivalent seismic coefficient (k) isused within a numerical slope stability calculation.
Since the seismic coefficient (k) designates thehorizontal force to be used in the stability analysis,
its selection is crucial. The coefficient is related tothe site peak ground acceleration components.

2. NARGES soil nailed wall
In this research, a soil nailed wall which is locatedin Tehran is selected as a case study. The soil
nailed wall has a depth of 23 m. The properties ofthe soil layers are presented in Table I. The model
used in analysis was Hardening model. Fig. 1
illustrates a profile of the soil nailed wall.
Table I. Geotechnical properties of excavation

The stability analysis was carried out by using
PLAXIS software which is based on the finite
element method. Recommendations for the757

minimum factor of safety for pull out resistanceand nail bar tensile strength are set at 2 and 1.8
respectively [1]. Table II presents Characteristics ofnails used in the designed excavation.

11

1.5

14

32

12

1.5

14

32

13

1.5

14

32

2.1 Hardening Model
The hardening soil model is an advanced model forthe simulation of soil behavior. In contrast to anelastic perfectly – plastic model, the yield surface
of a hardening plasticity model is not fixed inprincipal stress space, but it can expand due toplastic straining. Distinction can be made between
two main types of hardening, namely shear
hardening and compression hardening. Shear
hardening is used to model irreversible strains due
to primary deviatoric loading. Compression
hardening is used to model irreversible plasticstrains due to primary compression in oedometerloading and isotropic loading. Both types ofhardening are contained in the present model [2].

Fig. 1 Soil nailed wall with depth 23m

2.2 Damping

Geogrid and Beam elements were used to modelnails and facing elements, respectively. Input
parameter definitions in PLAXIS require averaging
the effect of a three-dimensional problem to a twodimensional problem. The simulation of the entire
soil-nail wall construction process was carried out
in a sequence of stages. In each construction stage
a sufficient number of calculation steps was used toobtain an equilibrium-state.

No.

Horizontal
distance(m)Length(m)Diameter
(mm)

1

2

12

28

Natural dynamic systems contain some degree of
damping of the vibration energy within the system;otherwise, the system would oscillate indefinitely
when subjected to driving forces. In soil and rock,natural damping is mainly hysteretic — i.e.,
independentoffrequency(Gemant
andJackson(1937) and Wegel and Walther (1935)).
Rayleigh damping was originally used in theanalysis of structures and elastic continua, to damp
the natural oscillation modes of the system. A
damping matrix, C, is used, with components
proportional to the mass (M) and stiffness (K)matrices:

2

2

12

28

C = α M +βK

3

1.5

14

32

4

1.5

14

32

5

1.5

14

32

6

1.5

14

32

where α = the mass-proportional damping constant;and β= the stiffness-proportional damping constant[2]. The damping ratio of soils generally varies
between 2 – 5%, and this value for the structuresvaries between 2 – 10% (Biggs 1969) [3]. Within
the scope of this study, the damping ratio used is
5%.

7

1.5

14

32

8

1.5

14

32

9

1.5

14

32

10

1.5

14

32

Table II. Characteristics of nails used in thedesigned excavation(1)

3. Problem Analysis
It is attempted to achieve a relation between groundseismic coefficient and ground motion amplitude toconclude the value of seismic coefficient andseismic study of this wall. The procedure ofanalysis is as follows:- Dominant frequency was obtained by thefollowing procedure. First of all different loading758

Therefore these bottom-row nails avoid from slope
failure by having anchoring effect and wheneverplastic surfaces propagate behind the whole nails, Itmeans that the bottom-row nails lost theirs
anchoring effect and the structure will fail. Failure
may occur at a specified pullout displacement ofbottom-row nails. Hereafter the mentioned pullingout displacement will be known as critical pullingout displacement[4].

frequency with a specific duration was applied tothe model. Amplitude of the vibration is a factor ofsystem response; so the vibration amplitude of eachselected loading frequency was extracted andfinally the Frequency corresponding to maximum
response of the system is the dominant frequencyof the structure. Usage of this method, wasconcluded to the natural frequency of 2hz for thiswall (see fig. 2).

This method is based on supposition that failure
occurs at the constant pullout displacement ofbottom-row nails for both cyclic and pseudo-static
analysis methods. Using this concept, relation
between the critical seismic coefficient and thecritical peak acceleration can be achieved. The wall
was induced by different pseudo static analysis toobtain critical seismic coefficient (a factor of safetyequal to one). Clearly, critical seismic coefficient isobtainable by a linear regression between seismic
coefficient with the factor of safety values. The
below diagram (Fig.3) presents safety factors
obtained for different seismic coefficient by pseudo
static analysis.

Fig 2: Reactions by different parts of wall todifferent frequencies- The wall was induced by different pseudo static
analysis to obtain critical seismic coefficient (afactor of safety equal to one). The k h is defined as
fraction of the normalized horizontal acceleration(A m ), which acts at the center of gravity of thewall-soil mass (AASHTO, 1996). A m is a functionof the normalized peak ground acceleration
coefficient (A), which is the actual peak groundacceleration normalized by the acceleration ofgravity (g), and is defined as[1]:Am =A (1.45 - A)

Fig. 3 Safety factors obtained for pseudo static
analysis against horizontal seismic coefficient
From the above data (Fig.3), critical seismic
coefficient (a factor of safety equal to one) byassuming linear change of variables and throughlinear regression would be achieved 0.165g. Plot ofpullout displacement for bottom-row nails versus
different seismic coefficients is presented in Fig. 4.

(2)

Assuming linearity of variable changes, the criticalpullout displacement was achieved.
- At last through cyclic loading the wall was
induced by different seismic amplitudes.
- By using plot of peak seismic amplitude versus
critical pullout displacement which has beenachieved through second step, the critical seismic
amplitude can be obtained.

4. Analysis Results
The maximum axial forces mobilize in the bottomrow nails with increasing the dynamic forces.

759

k h /A m =0.16/0.27=0.59
R

R

R

R

- Results of loading with frequency of 2hz and 8
cycles is expressed as follows (see Fig 6):

Fig 4: displacements caused by pseudo static
analysis versus corresponding seismic coefficients
Assuming linearity of variable changes, based on
the above diagram (Fig.4), critical displacement for
critical seismic coefficient of 0.165g was obtained
8.3cm. From the above results, the critical pullout
displacement achieved. Through cyclic loading thewall was induced by different seismic amplitudes
and different number of cycles.

Fig. 6 displacement caused by seismic loading on
the wall for 8 cycles, frequency of 2hz anddifferent amplitude
As illustrated in the Fig.6 maximum seismic
acceleration was obtained 0.18g. This amount iscomparable with 0.23g for normalized seismic
acceleration (Am).

- Results of loading with frequency of 2hz and 5
cycles on the soil nailed wall is illustrated as below(see Fig.5) :Am=A*(1.45 –A)
k h /A m =0.16/0.23=0.69
R

R

R

R

- Results of loading with frequency of 2hz for 10
cycles is as follows:

Fig. 5 displacements were conducted by seismic
loading on wall for 5 cycles, frequency of 2hz anddifferent peak amplitudes
By using the plot of peak seismic amplitude versus
critical pullout displacement and the estimate ofseismic amplitude at the critical pullout
displacement, the critical seismic amplitude can beobtained (see Fig 5). Based on the Fig.5, maximum
seismic acceleration (A) was obtained 0.22g. Thisvalue equals to 0.27g for normalized seismic
acceleration (Am).
Am=A*(1.45 –A)

Fig. 7 displacement caused by seismic loading on
the wall for 10 cycles, frequency of 2hz anddifferent amplitudes
As illustrated in the Fig.7, maximum seismic
acceleration was obtained 0.172g. This amount iscomparable with 0.22g for normalized seismic
acceleration (Am).

(2)

Am=A*(1.45 –A)

760

kh/Am=0.16/0.22=0.72
The rate of critical seismic coefficient tonormalized critical seismic coefficient, as shown by
the results (see Fig 5 to 7), is a limited range
between 0.59 to 0.72.
5.Conclusion
The results of seismic analysis of soil nailed wall in
the project of NARGES have been reported. Twonumerical analysis methods are applied to simulate
the seismic behavior. The computational results areas below:
The values of Am depend on the regional tectonic
setting and are obtained from seismic maps, for the
project region this amount is 0.35g. Obtained
results from this method indicates that seismic
coefficient is in form of a limited range of groundseismic normalized horizontal acceleration(i.e.,0.59 - 0.72). It means that the seismic
coefficient is in form of a range (0.2g to 0.25g)
which it is higher than critical seismic coefficient(0.165g).
The comparison of these values with the criticalseismic coefficient indicates that this wall isprobably vulnerable under earthquake effects, so its
seismic resistance should be increased.
6. References1.

FederalHighway Administration
Geotechnical Engineering CIRCULAR
NO.7 Soil Nail Walls" prepared for us
Department of Transportation , FHWA 0-IF03-017, 2003

2.

Plaxis Manual , Version 8.2 , A.A.Balkema ,Netherland , 1998

3.

Itasca. 2000. “FLAC (Fast Largrangian
Analysis of Continua) User's Manuals”.
Minneapolis: Itasca Consulting Group,Inc.

4.

Komak panah, A. and Majidian, S.
“Assessment of soil nailed excavations
seismic failure Under cyclic loading andpseudo-staticforces.
International
conference on Recent advances ingeotechnical earthquake engineering andsoil dynamics,2010761

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Purification System of Ocean Sludge by Using Coagulants
and Activating Microorganisms
Kyoichi OKAMOTO 1 and Kenji HOTTA 1
CST, Nihon University, Japan
ABSTRACT
Ocean sludge exerts a very big environmental load to local sea area. Here, attention was paid to micro-bubble
technology for application to the purification of the sludge. The important point in this technique is to activate thebacteria existing in the area by micro-bubbles. We had developed a method for decomposing the sludge by using of
microorganisms in an aerobic state by micro-bubble. We had very good results by the method, and also understood thathydrogen sulfide is reduced at first and then the nutrients will be reduced. Here, we have also a technique forpurification by using “coagulants”. Therefore, we proposed the experimental way which hydrogen sulfide is reduced at
first by using “coagulants” and then the nutrients will be reduced by micro-bubble and activating microorganisms.
Here, we used the detergents including enzyme as the microorganism activator. Our research object in this paper is tocheck the purification performance for sludge by our proposed experimental system. From the results of ourexperiments, we succeeded in reducing the time needed to purify the sludge, and we obtained the very good
performance.
Keywords: Coagulants Detergent Micro bubble Microorganism Activator Purification Sludge
1. INTRODUCTION
H2S [mg/l]

T-N
H2S

T-N [mg/l]

It is very important to reduce sedimentary sludge in the ocean.
Plans to reduce the sludge are usually dreading or sandcovering. Dredging is a simple way and aims to cut off thesludge. But after cutting off, treating the dredged sludge takes
much more time and, of course, cost. Sand covering, in
general, gives a big load to living organisms and theecological system. Here, a more efficient way is needed toreduce the sludge while not imparting environmental load in
the local sea area.
Now, we have micro-bubble technology. Micro-bubbles (thatis MB) can change conditions into an aerobic state. So, we
had developed a method for decomposing the sludge by usingof microorganisms in an aerobic state by micro-bubble. We
had very good results by the method in [1], [2]; for example, it
could be reduced the treatment days to 5 days. We had also
understood that hydrogen sulfide is reduced at first in an
aerobic state by micro-bubble and then the nutrients will be
reduced, shown in Fig. 1.
Here, we have also a technique for purification by using“coagulants”. We had also very good results by using
coagulants in [3].
Therefore, we propose the way which hydrogen sulfide is
reduced at first by using coagulants and then the nutrients will
be reduced in an aerobic state by micro-bubble and activating
microorganisms, since we are going to reduce the treatmentdays more.
We used the detergents including enzyme as the
microorganism activator, since we can obtain it easily andalso put on the market, in [4].

Fig.1 General results of H2S and T-N by our experiments
Our research object in this paper is to check the purification
performance for sludge by our proposed experimental system.
2. EXPERIMENTAL SYESTEM
Our proposed method for purification experiment has 2 steps.
First step is that a hydrogen sulfide is reduced at first by using
coagulants. Second step is that the nutrients will be reduced in
an aerobic state by micro-bubble and activating
microorganisms.
2.1 Experimental Apparatus and Procedure in Step 1
Sludge and seawater were put in the tank (Long70 xWidth47
xHeight 28cm) and then were mixed by a water pump
(300litter/hour) for 10 minutes after coagulants were put in,
shown in Fig.2.

762

Flocculated Sludge

Fig. 6 Mechanism of Coagulants
Mixing Tank
Remaining Seawater

Fig.2 Schematic Representation of Experimental Apparatus
in Step 1
Here, seawater is 90 (litter) and ocean sludge is 3 (kg) whichwere picked up from Funabashi Port in Tokyo in Japan, as
shown in Fig.3 and 4. Concentration of coagulants was 400(ppm). After sludge was changed into flocculated situation,flocculated sludge will be used for reuse and the remaining
seawater is used for next experiment in step 2.

2.2 Coagulants in this Paper
Our used coagulants are neutral and inorganic substance,
shown in Fig.5. This is neutral and inorganic and the main
ingredients of this are compound alumina and silicate. When
we use this, we have to mix strongly with water after putting
in. This has a characteristic of adsorption and fix-separation
to cohere not only nasty substance of floating but also
dissolving. This has also a characteristic of good desiccate
and fixing heavy metal stable. Reaction mechanism of thecoagulants is shown in Fig.6, in [5].
2.3 Experimental Apparatus in Step 2
The experimental devices in step 2 consist of two parts,
shown in Fig. 7 and 8. The water circulates through two tanks.
In one tank (Width40xLength28xHight28cm), micro-bubbles
are generated. The micro-bubbles have micro-size diameterand high solubility. This means the water with highconcentration of dissolved oxygen circulates through thesetanks. The other part is the experimental tank (W60xL29
xH35cm). The remaining seawater in step 1of our experimentis put in this tank. Here, a micro-bubble generator is based on[6] and the flow rate is 1200 (litter/hour). The flow rate ofwater pumps connected each tanks are 300 (litter/hour).

Fig.3 Gathering Instrument for Sludge

Tokyo

Tokyo Bay
Yokohama

Circulation
YokosukaTokyo BayMB
MBGenerator
発生装置
Water Pump
撹拌用ポンプMB Generating Tank
Experimental Tank

Funabashi Port

Fig.4 Catching Point of Sludge at Funabashi Port in Tokyo
Bay

Fig. 5 Coagulants in this Research

Fig. 7 Schematic Representation of Experimental Apparatus
in Step 2

Fig. 8 Picture of Experimental Apparatus in Step 2763

2.4 Experimental Procedure in Step 2
After creation of flocculated sludge by experiment as Step 1,
the remaining seawater was put in the tank, and then the
micro-bubble device and the water pumps were powered on.
Here, room temperature was set 25 degrees Celsius by an air
conditioner. After 6 hour, the microorganism activator was
put in the experimental tank.
In this experiment, one kind of detergent including enzyme as
the microorganism activator is used by the reasons denoted
the best results in [4]. Including enzyme in the detergent are
mainly lipase, protease and amylase. Amount of the detergent
as experimental conditions are shown in Table. 1. The
appropriate weight of this detergent is 10 (gram). For
comparison, there is one more case by using the
microorganism activator which had been getting effective
results in purification for grease trap at gas station, written by[2], so this case is named “Case 4, 2008”.
The amount of sludge and seawater is 1(kg) and 30(litter), peran experimental case.
Dissolved oxygen (DO), water temperature and pH are
measured by using of multi-parameter water quality meter.
Ammonium nitrogen (NH 4 -N), total nitrogen (T-N) and total
phosphorus (T-P) are measured by using ofdigital-water-analyzer by digital “Packtest”, by water filteredafter sampling in experimental tank.
Table. 1 Amount of Detergents as Experimental Conditions
Case

Amount(Times)(Additional / Appropriate)

①
②
③

5
6
7

④

Reference by Okamoto et al. in 2008

Table. 2 Results on Experiment of Step 1
Temp. [deg. Celsius]
pHDO[mg/l]
H2S[mg/l]
NH4-N[mg/l]
T-N[mg/l]

Before
22.32
7.37
0.70
0.02
3.42
2.6

After
23.80
7.95
6.82
0.00
2.38
1.4

3.2 Effects on Experiment of Step 2(1) Effects on the density of sulfate and the contents of sulfur.
Checking the mechanism in this experimental system, thedensity of sulfate was analyzed in Case 2. The density ofsulfate is shown in Fig. 8, adding the results by [2] in 2008
which is Case 4. The results were obtained by the ironchromatography. It seems that sulfate increases due to theactivity of sulfur bacteria in Case 4. But the density of sulfate
in Case 2 doesn’t increase. It seems the reduction of H 2 S
became by lower density of sulfate, and microorganisms also
were reduced a little by experiment of step 1.
The content of sulfur is shown in Fig.9, adding the results by[2] in 2008 which is Case 4. The results were obtained by theelementary analysis. Especially in Case 4, the content ofsulfur up to 6 hours did not change but after 6 hours it
decreased rapidly since the activator of microorganisms was
put in the experimental tank after 6 hours. But the content ofsulfur in Case 2 is almost zero. It seems that microorganisms
changed the sludge into the sulfur by the ordinary method, but
the present method could not change the sludge into thesulfur.

3 RESULTS AND CONSIDERATION
Fig. 8 Results of the Density of Sulfate in Step 2

3.1 Effects on Experiment of Step 1

1(%)

Results of the experiment of step 1 by using coagulants areshown in Table. 2. The results on water temperature and pH
are a little change, compared with each other. We can see thesituation changed into an aerobic state from the results on DO
by mixing sludge with coagulant. We can also see the resultsof H 2 S were decreased zero by the experiment of Step 1. On
the other hand, the results on the nutrients; NH 4 -N and T-N
also decrease to 70 and 53%.

②
④

0.5

0
0

24

48
72
Time [hour]

96

120

Fig.9 Results of the Content of Sulfur in Step 2764

(2)Effects on H 2 S.
Measured results of H 2 S in all cases are shown in Fig.10. The
result in Case 4 is a usual tendency. On the other hand, theresults in Case 1, 2 and 3 are almost zero. It seemed
coagulants in the experiment of Step 1 reduced H 2 S. From theresults, our method is very good.
(3)Water Temperature, pH and DO as Environmental
Conditions for Experiment of Step 2.
Measured results of water temperature, pH and DO in all
experimental cases are shown in Fig.11 to Fig.13,
respectively. Water temperature up to 24 hours increased
rapidly and then became constant about 25 to 30 degree
Celsius. It seems this was caused by the heat from friction of
the micro-bubble device. The tendency of pH has up anddown in Case 1, 2 and 3, but became constant about 7.6 to 8.5.
It seemed it is better to use for experiment. Dissolved oxygen(DO) in Case 4 up to 12 hours increased and then became
constant. It seems oxygen dissolved and reached thesaturation point. Case 1, 2 and 3 have a tendency of up anddown but these have enough oxygen by experiment of Step 1.
From these results, we can recognize anaerobic statechanging into aerobic state. It seems that these results are very
good conditions for growth of microorganisms.

H2S[mg/l]

0.05
0.04
0.03

Fig.12 Changes in pH as Environmental Conditions byExperiment of Step 2① 5 times② 6 times③ 7 times④ 2008

0.02
0.01
0

Fig.13 Changes in DO as Environmental Conditions byExperiment of Step 2
12 24 36 48 60 72 84 96 108 120
Time[hour](4)Effects on NH 4 -N, T-N and T-P.
Measured results of NH 4 -N in Cases 1, 2 and 3 are shown in
Fig.14. The results of NH 4 -N in Case 1 and 2 decrease andbecome zero until 24 hours. But the result after 84 hours inCase 1 and 3 increase a little. It seemed the Case 2 has very
good result, compared with Case 1 but Case 1 and 3 become
no effect after 84 hours. Measured results of T-N and T-P in
all cases are shown in Fig.15 and Fig.16. The results of T-N
have a tendency of decrease until 84 hours. Case 2 has very
good results that T-N is purified until 24 hours. On the other
hand, the results in Case 1, 2 and 3 after 96 hours increase
unexpectedly. It seemed Case 2 has very good performance
for T-N, but the results are not effective after 96 hours.
The results of T-P in Case 1 and 2 have good performance,compared with Case 4 by results in 2008. The results of T-P
in Case 2 become no effect after 96 hours. From thisexperiment, we pointed out the decrease on T-N

Fig.10 Changes in H 2 S by Experiment of Step 2

Fig.11 Changes in Water Temperature as Environmental
Conditions by Experiment of Step 2765

NH4-N [mg/l]

4.1 Re-Experiments
Experimental conditions for the method of adding theactivator twice are shown in Table. 3. Second adding amountof Case 5, 6, 7 is 2, 4, 6 times of activator at time=60hours,
respectively. For comparing the effects, the results of Case 2
in experiment of step 2 are used. The other conditions arealmost same of experimental of step 2. Measured results onenvironmental condition of this re-experiment were also
almost same of experiment of step 2.

4.2 Effects on NH4-N and T-N
Measured results of NH 4 -N in Cases 5, 6 and 7 are shownin Fig.17, with Case 2 in experiment of step 2. The results of
T-N are also shown in Fig.18.
The re-experimental system has very good performancefrom these results. NH 4 -N of Case 5 can be purified until 72
fours. T-N of Case 5 and 6 can be also purified until 84 hours.
It seems that we could get very good performance forpurification by adding the activator of microorganisms andreduce the time for purification, since purification
performance was shortage from the results in experiment ofstep 2.

Fig. 14 Changes in NH 4 -N for Experiment of Step 2

Table. 3 Method of Adding Activator as Re-Experimental
Conditions

Fig. 15 Changes in T-N for Experiment of Step 2

Case

⑤
⑥
⑦
②

NH4-N[mg/l]

4.0

Fig. 16 Changes in T-P for Experiment of Step 2

4 RSULTS OF RE-EXPERIMENT ANDCONSIDERATION
It seems our proposed method is very good until 84 hours
from the results of experiment of Step 2. For getting the best
purification, we had carried out re-experiments to put theactivator twice in experiment of step 2, because the results are
not effective after 96 hours.

3.0

Method of Adding Activator(Amount(Times)=Additional/Appropriate)
First adding
Second adding
at time=0
at time=60hour

2 times
4 times
6 times0

6 times⑤6+2times⑥6+4times⑦6+6times②6times

2.0
1.0
0.0
0 12 24 36 48 60 72 84 96 108 120
Time[hour]

Fig.16 Results on NH 4 -N by Re-Experiment766

Fig.17 Results on NH4-N by Re-Experiment6 ACKNOWLEDGMENT
7.0
T-N[mg/l]

6.0
5.0
4.0

The authors would like to express sincere thanks to Dr. T.
Toyama, Nihon University, in Japan

⑤6+2times⑥6+4times⑦6+6times②6times

7 REFERENCES

3.0
2.0
1.0
0.0
0 12 24 36 48 60 72 84 96 108 120
Time[hour]

Fig.17Results
Results on
on T-N
Fig.18
T-Nby
byRe-Experiment
Re-Experiment

5 CONCLUSIONSWe proposed the system which hydrogen sulfide is reduced at
first by using coagulants and then the nutrients is reduced in
an aerobic state by micro-bubble and activating
microorganisms. Here, we used the detergents including
enzyme as the microorganism activator.
We also carried out purification experiments and thenchecked the purification performance for sludge by our
proposed experimental system. From the results by theexperiment of Step 1 and 2,
1) It seemed coagulants in the experiment of Step 1 reduced
H 2 S.
2) It seemed microorganisms also are reduced a little bycoagulants in the experiment of step 1.
3) The performance of purification in Case 2 is very good forthe results on T-N and also NH 4 -N, until 84 hours.
4) The results of T-P in Case 1 and 2 have good performance,compared with the results in 2008.
Here, we carried out re-experiment which was put twice ofactivator at 0 and 60 hours, since the purification performance
was shortage from the results in experiment of step 2.
From the results by re-experiment, we could get very good
performance for purification and reduced the time forpurification.
Finally, it is seemed our proposed experimental system isvery good.

[1] Okamoto K, Hotta K, Toyama T, and Kohno H,“Experiments on purification of ocean sludge byactivating microorganisms,” Proc. of ISOPE-2011-TPC-574.
[2] Okamoto K, and Hotta K, “Purification experiments on
sedimentary sludge by microorganism activator,” Proc.
of Pacific Congress on Marine Science and Technology(PACON2008).
[3] Okamoto K, and Hotta K, “Experiment on purification forwater quality by using of cohesion powder,” Proc. of
ISOPE-2006-JSC-306.
[4] Okamoto K, Hotta K, Toyama T, and Kohno H,“Purification experiments of ocean sludge by activating
microorganisms -by using detergent including enzyme as
the activator-,” Proc. of ISOPE-2012-TPC-379.
[5] Matsui R, Okamoto K and Hotta K, "Water Purification
Experiments by Micro Bubble," Book of Recent
Advances in Marine Science and Technology, 2006,
pp.119-126.
[6] Imai T, Shioshige K, Ukita M, Sekine M, Higuchi T,
Fukagawa K and Fujisato T, “Development of Device for
Dissolving High Concentration Gas for Purification atbottom layer around enclosed water area like Dam
Lake,” Proc. 39th Forum on Environmental Engineeringof Civil Engineering Society, 2002, 10-12.
[7] Matsuo K, Maeda, K, Ohnari, H, Tsunami, Y, and Ohnari,
H, “Water Purification of a Dam Lake Using Micro
Bubble Technology,” Progress in Multiphase Flow
Research I, 2006, pp.279-286[8] Mizoguchi M, “Observation of Water Purification Process
in a Moat by Microbiological Treatment Technique,” Proc.
of Biological Resource in Mie University, No.16, 1996,
pp.25-37.
[9] Hibino T and Matsumoto H, “Distribution of Fluid MudLayer in Hiroshima Bay and its Seasonal Variation”, Proc.
Civil Engineering, Vol. 62, No 4, 2006, pp.348-359.

767

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Mechanical properties of volcanic products mixed with industrial wastes
1

K. Yamamoto1, T. Negami2, M. Hira1, N. Aramaki3 and Y. Hayashi4Kagoshima University, 2Saga University, 3Sojo University and 4Kyushu Sangyo University, Japan

ABSTRACT
Mt. Shinmoedake, located on the boundary between Kagoshima and Miyazaki has erupted on January 26, 2011 after aninterval of 189 years. The Volcanic products are estimated to be a maximum of 40-80 million tons. On the other hand,the frequency of eruptions at Mt. Sakurajima in Kagoshima peaked in 2011, the highest since observations began in
1955. Large amounts of volcanic ashes from Mt. Sakurajima were falling in the surrounding areas. Most of volcanic
products have been disposed as industrial wastes, but the supply of these volcanic products is anticipated to be unending
for the foreseeable future. The purpose of this study is to contribute to the formation of a recycling-oriented society in
the Southern Kyushu area by utilizing the volcanic products as local natural resources. This study investigates themechanical properties of these volcanic products mixed with industrial wastes for environmental protection.
Keywords: volcanic products, compaction, California bearing ratio, recycling-oriented society

1. INTRODUCTION
Mt. Shinmoedake and Mt. Sakurajima are located in Southern
Kyushu and are both active volcanoes in Japan. Due to the recent
serious seismic activities in Japan, these eruptions have been
recently activated. A large amount of volcanic products are
already supplied in the Southern Kyushu area and the constant
supply of these products is anticipated from now on. The purposeof this study is to contribute to the formation of recycling-oriented
society by utilizing the volcanic products as local natural
resources. This study investigates the mechanical properties of the
volcanic products mixed with industrial wastes for environmental
protection. The mixture of industrial wastes is scrapped ceramics,
reclaimed (recycled) gypsum and slaked lime, etc. The mixture
ratio is the weight ratio for the natural water content of the
volcanic products. In this paper, the possibility of the effective use
as a frictional geomaterial ( φ materials) is considered.

Figure 3 shows the grain size distribution curve. The
volcanic products of Shinmoedake were mainly sand, but thecontent (by percentage) of fines was higher for material from
Sakurajima. Thus, the mixture of each volcanic product was
taken into account, because the properties of pH and grain
size distribution were different. The mixture ratio was set as
the weight ratio 70:30, 50:50, 30:70 (%) for the natural watercontent of each volcanic product. When the mixture ratio of
Shinmoedake increases, the percentage content of fines isreduced. Next, Figure 4 shows the compaction curve for eachvolcanic product. The compaction test [1] was conducted by
A-a method, where the compaction energy is low and thesample is repeatedly used. It is found that the volcanic
products of Shinmoedake show a higher optimum watercontent and lower maximum dry density than those of
Sakurajima.
3. RESULTS AND DISCUSSION

2. PHYSICO-CHEMICAL PROPERTIES OF
VOLCANIC PRODUCTS
The volcanic products of Mt. Shinmoedake and Mt.
Sakurajima sampled in Miyakonojyo and Kagoshima, as
shown in Figure 1. Figure 2 shows the microscopic
observation of each volcanic product. It can be observed thatthe particle size of Shinmoedake is obviously larger than thatof Sakurajima. Table 1 shows the physico-chemical
properties of each volcanic product. The volcanic products of
Sakurajima showed heavier density and lower permeability
than those of Shinmoedake. In addition, those of
Shinmoedake and Sakurajima were found to have weakacidity and alkaline, respectively.

The California bearing ratio (CBR) test [2], which is available
for the evaluation of the strength of subgrade or subbase wascarried out in the laboratory to obtain the design CBR value.
Slaked lime, reclaimed gypsum and scrapped ceramics areused as the mixture of industrial wastes, and the mixture ratiois the weight ratio for the natural water content of eachvolcanic product. Figure 5 shows the loadingpressure-penetration curve for each artificial geomaterial.
Figures 5(a) and (b) are the cases of Shinmoedake andSakurajima mixed with slaked lime. The loading pressure of
mixtures of Shinmoedake is higher than that of only
Shinmoedake, when the mixture ratio of slaked lime increases.
On the other hand, Sakurajima mixed with slaked lime atratios of 10% and 20% can not exceed the loading pressure of
only Sakurajima. This is because the volcanic products of768

(b) Sakurajima(a) Shinmoedake

Figure 1 Volcanic products sampled in southern Kyushu area.

(a) Shinmoedake(b) Sakurajima

Figure 2 Microscopic observation of volcanic products.
Table 1. Physico-chemical properties of volcanic products

Natural water content (%)

Shinmoedake

Sakurajima

10.5

10.5

3

2.685

2.708

Minimum density (g/cm )

1.126

1.385

Soil particle density (g/cm )3

Maximum density (g/cm3 )Permeability coefficient (cm/s)
pH

1.460

1.654
-2

1.6×10
6.40

Figure 3. Grain size distribution curve.
769

2.0×10-4
7.91

ρ d max = 1.36 g / cm3
wopt = 23.8%

ρ d max = 1.63g / cm3
wopt = 15.2%(a) Shinmoedake

Figure 4 Compaction curve.

(b) Sakurajima(a) Shinmoedake plus slaked lime(b) Sakurajima plus slaked lime(c) Shinmoedake plus Sakurajima(d) Sakurajima plus scrapped ceramics

Figure 5 Loading pressure-penetration curve.
Shinmoedake are mainly sand with weak acidity and those of
Sakurajima are a finer sand and alkalinity.
Figure 5(c) shows the case of the mixture of each volcanic
product (Shinmoedake and Sakurajima). The obvious
difference in the loading pressure between only Shinmoedakeand Sakurajima is seen and the loading pressure-penetration
curves of the mixture of each volcanic product are caught
between the curves of only Shinmoedake and Sakurajima.
The loading pressure of Sakurajima is clearly higher than thatof Shinmoedake. Additionally, we have confirmed that theloading pressure is somewhat dependent on the mixture ofsamples by the inside observation of samples after the
penetration.
Figure 5(d) is the case of Sakurajima mixed with scrapped
ceramics. When the mixture ratio of scrapped ceramics
increases, the loading pressure tends to become higher. But
the mixture of scrapped ceramics to Sakurajima has no effectbefore the penetration reaches 5 mm. The trend was the samefor Shinmoedake. It is thought that the interlocking between
these volcanic products and scrapped ceramics would not be

770

Table 2. CBR (%) obtained from volcanic products and those mixture
Shinmoedake

Sakurajima

only volcanic products

11.59

25.86

volcanic products plus slaked lime 10%

21.04

17.30

volcanic products plus slaked lime 20%

33.88

21.40

volcanic products plus slaked lime 30%

44.40

59.74

volcanic products plus reclaimed gypsum 10%

28.09

51.07

volcanic products plus reclaimed gypsum 20%

42.65

113.43

volcanic products (70%) plus scrapped ceramics (30%)

5.44

14.89

volcanic products (50%) plus scrapped ceramics (50%)

4.28

21.22

volcanic products (30%) plus scrapped ceramics (70%)

-

21.31

MixtureSakurajima (30%) plus Shinmoedake (70%)

27.64

Sakurajima (50%) plus Shinmoedake (50%)
Sakurajima (70%) plus Shinmoedake (30%)

22.29
19.62

effective. Note that there was no expansion in all cases shownin Figure 5. Furthermore, the mixture of reclaimed gypsum toeach volcanic product is more effective than that of slaked
lime for the loading pressure-penetration curve. Finally, thedesign CBR(%) values from volcanic products and thesemixtures are summarized in Table 2. The volcanic products ofShinmoedake mixed with slaked lime 10% and 20% show
higher CBR values than those of Sakurajima. In other allcases, the tendency was the reverse (the mixtures of
Sakurajima are almost always a higher loading pressure thanthose of Shinmoedake).

5. REFERENCES[1] Japanese Geotechnical Society, Soil test –basis andguideline- (first revised edition): Maruzen Print Co.
Ltd., 2011, pp. 71-78 (in Japanese).
[2] Japanese Geotechnical Society, Soil test –basis andguideline- (first revised edition): Maruzen Print Co.
Ltd., 2011, pp. 79-90 (in Japanese).

4. CONCLUSIONS
In general, it would be valid to utilize these volcanic products
as frictional geomaterials under sufficient compaction control.
Also, there are no problems in using each volcanic product asa subgrade soil because the design CBR values of volcanic
products only are greater than 10% with no expansion. The
mixture of slaked lime and reclaimed gypsum were effective
for the volcanic products of Shinmoedake, but the mixture ofreclaimed gypsum was more effective than that of slaked lime
for those of Sakurajima. Furthermore, since the volcanic
products of Sakurajima are mainly finer sand and very little
spread, the strength of the mixtures of Sakurajima tends tovary a little than those of Shinmoedake.
For future work, it will be possible to consider thedevelopment of base materials for fishing banks and seaweed
beds as a practical use of these volcanic products andindustrial wastes.

771

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Measurement of Hydraulic Conductivity for Peat Ground Using CPTU
Hirochika Hayashi and Satoshi Nishimoto
Civil Engineering Research Institute for Cold Region, Japan

ABSTRACT
In-situ permeability tests were conducted on peaty ground in Hokkaido, Japan to clarify the permeability of such groundand a method for determining permeability parameters for related finite element (FE) analysis. As a result, it was
confirmed that the hydraulic conductivity (coefficient of permeability ki) obtained from in-situ permeability testing
which presents the mean permeability of the peat layer ranged from 1.7×10-7 to 1.1×10-5 (m/s). In addition, adissipation test using the CPTU was conducted and time values for 50% dissipation of excess pore water pressure (t50)were measured. It was found that the coefficient of permeability ki of peat decreased with an increase in the t50.
Keywords: peat, coefficient of permeability, in-situ permeability test, CPTU, oedometer test

1. INTRODUCTION
Peat ground, which consists of very soft and problematic soil,
is widely distributed in Hokkaido, Japan. When constructing
embankments on soft ground types such as peat, considerable
ground deformation occurs caused by both the consolidationand shear (Fig. 1). A highly effective method of analyzing
such ground deformation is soil/water coupled finite element
analysis (FE analysis) using a constitutive model that can take
both consolidation and shear into account. Hayashi et al. [1]clarified that FE analysis using Cam clay model (Schofield
and Wroth [2]) was effective in analyzing peat ground. It cantherefore be understood that FE analysis using anelasto-plastic or visco-plastic model for clay is a suitable
method for clarifying the deformation of peat ground.
Peat ground has many engineering properties that have
not yet been clarified, which poses the practical problem of
how to determine the soil parameters for analysis. Hayashi etal. [3] proposed a method of determining parameters forelasto-plastic FE analysis of peat, but merely reached atentative conclusion in terms of hydraulic conductivity(coefficient of permeability) determination. Therefore, the2. INVESTIGATION SITES
The permeability characteristics of peat ground were
investigated at nine sites (Mihara and Shinotsu in Ebetsu City,
Tsuruno in Kushiro City, Warabidai in Toubetsu Town,
Riyamunai and Kami-Riyamunai in Kyowa Town, Taura in
Mukawa Town, Onobunai in Teshio Town and Asajino in
Sarufutsu Village) in Japan’s Hokkaido area.
The natural water content (Wn) of the peat studied ranged
from 240 to 1,023%, and ignition loss (Li) was 23 to 95%.
The study was conducted for peat with a wide variety ofengineering properties, from the mostly decomposed
low-water-content type to the high-water-content fibrous type.
For comparison with peat, organic clay and clay underlying
the peat ground were also tested.
3. TEST PROCEDURES
An in-situ permeability examination involving a borehole anddissipation of pore water pressure using the electric cone
penetration test (CPTU) method was conducted. In thelaboratory, an oedometer test was conducted with undisturbed
samples, and the results were used to determine coefficientsof permeability. The test methods are detailed below.

embankmentconsolidation

coefficient of permeability and changes in this value resulting
from consolidation for peat ground were investigated basedon the results of an in-situ permeability test and oedometer
test, and influences related to differences in the coefficient inpeat ground on the results of FE analysis were studied.

shear

peat

3.1 In-situ permeability test
The in-situ permeability test was conducted in accordancewith the Method for Determination of Hydraulic Properties of
Aquifer in Single Borehole (JGS 1314-2003) in the Japanese
Geotechnical Society standards (Fig. 2).
The in-situ
coefficients of permeability were found using Eq. (1), in

sandFig. 1 Typical deformation of peat ground caused byembankment loading772

which ki is the on-site coefficient of permeability (m/s), d isthe diameter of the water level measurement section (m), L is
the length of the test section (m), D is the diameter of the test
section (m), m is the slope of the straight section of the log s-t
curve found from water level measurement, s is the difference
between the steady and measured water levels (m), and t is theelapsed time (s).
ki = 0.66 d2 / L log (2L / D) m

Water- level gauge(1)

3.2 Dissipation test using the CPTU method
In the CPTU method, pore water pressure is generated when acone penetrates the ground. If the ground is left untouched
after penetration stops, the pore water pressure generated
dissipates over time, and the time it takes to do so is closely
related to ground permeability. Dissipation testing of porewater pressure using the CPTU method involves estimation
for the coefficients of ground permeability and consolidation
based on this phenomenon. High applicability of the test forclay ground has been reported previously (e.g., Robertson etal. [4]; Tanaka et al. [5]).
Figure 3 shows the shape of the cone used in this study.
It had a base area of 1,000 mm2 and a point angle of 60°.
The filter for pore water pressure measurement was placed
19.0 mm behind the base of the cone, as shown in Fig. 3. The
filter was ceramic and saturated with deaired water. The cone
was used to penetrate the ground with these specifications at a
rate of 20 mm/s, and penetration was stopped at the depth
where the in-situ permeability test was conducted. Changes
in pore water pressure were subsequently measured with thecone in place. To avoid interference between the in-situ
permeability test and CPTU application, the test sections
were positioned at least 2 m apart.
3.3 Oedometer test
The oedometer test was conducted in accordance with theJapanese Geotechnical Society standard on undisturbed
samples (dimensions: 60 mm in diameter and 20 mm inthickness) collected using a thin-walled tube sampler with afixed piston. As the effective overburden pressure of peat isvery small, consolidation pressure at the first stage was set to
4.9 kN/m2, which was half the standard value. The laboratory
coefficient of permeability was found using Eq. (2), where kl
is the laboratory coefficient of permeability (m/s), cv is thecoefficient of consolidation (cm2/d), mv is the coefficient ofvolume compressibility (m2/kN) and w is the unit weight ofwater (= 9.81 kN/m3).
kl = cv mv w / (8.64 × 108)

casing pipe(2)

4. COEFFICIENT OF PERMEABILITY OBTAINEDFROM IN-SITU PERMEABILITY TESTS AND
OEDOMETER TESTS
Figure 4 shows the relationship between the natural void ratio

773

test section L

Fig. 2 In-situ permeability testing method

Fig. 3 Shape of CPTU
and the field coefficient of permeability. Permeability
coefficient of peat, which ranged from 1.7×10-7 to 1.1×10-5(m/s), was greater by one or two orders of magnitude thanthose of clay and organic clay. The field permeability
coefficient of peat increased almost linearly on thelogarithmic axis with higher values of the natural void ratio.
The natural void ratio of peat approximately ranged from 8 to
15. These values were much larger than those for ordinary
soils, indicating that peat is predominantly composed of voids.
It should be noted, however, that the coefficient of
permeability for peat is smaller than that for sand, which has alower natural void ratio. Ohira [6] pointed out that most voids
in peat were enclosed in plant remains and only
approximately 20% of all voids were effective as
macroscopic pore water flow paths. Kogure [7] reported thatwater in peat contained humus formed as organic matterdecomposition, and was there-fore higher in viscosity thanthat found in ordinary soil. These factors can support that thecoefficient of permeability in peat shows middle value
between that of clay and sand despite of its very high natural
void ratio.
The relationship between the field and laboratory
coefficients of permeability is shown in Fig. 5. The

This section describes estimation of the in-situ coefficient of
permeability in peat ground based on dissipation testing ofpore water pressure using the CPTU method. Figure 7 shows
the time history of pore water pressure generated by cone
penetration (for the Riyamunai site at a depth of 3.0 m) as a
typical measurement result. The pressure reached an almost
steady state approximately 1,000 seconds after penetration
stopped, and showed a value of 28.1 kPa thereafter. The
result of pore water pressure estimation after themeasurement period (until 10,000 seconds) using thehyperbola method also confirmed that the steady waterpressure (Uf) value was 28 kPa. As the groundwater level ofthis site was 0.26 m below the surface, the static waterpressure at the depth where the dissipation test was conducted(G.L. -3.0 m) was 27.4 kPa, which was almost the same as theUf value. Similar results were obtained at other sites.
Past studies (e.g., Robertson et al. [4]; Tanaka et al. [7])used the time taken for 50% of pore water pressure todissipate (t50) in the dissipation curve (Fig. 7) as an index torepresent the dissipation status. Figure 8 shows therelationship between t50 and the in-situ coefficient of
permeability described in Section 5 based on a double
logarithm. There is a proportional relationship between thetwo, and the in-situ coefficient of permeability decreases
linearly with higher values of t50. Robertson et al. [7] also
reported that the coefficient of permeability fell withincreases in t50 based on examination of the relationshipbetween these values for clay.
The relationship shown in Fig. 8 is approximated using
Eq. (3), where ki is the in-situ coefficient of permeability(m/s) and t50 is the time taken for 50% of the pore water774

in-situ coefficient of permeability k i (m/s)

k i = 7.6 x 10-9 x exp(0.4 e 0)
1.0E-05

1.0E-06

1.0E-07peat
organic clay
clay
sand

1.0E-08

1.0E-09
0

5

10

15

20

in-situ void ratio e 0

Fig. 4 The relationship between in-situ void ratio andin-situ coefficient of permeability
in-situ coefficient of permeability k i (m/s)

5. IN-SITU PERMEABILITY ESTIMATION FROMDISSIPATION TESTING USING THE CPTU

1.0E-04

1.0E-04

k i /k l = 10
k i /k l = 100

1.0E-05
k i /k l = 1

1.0E-06

1.0E-07peat
organic clay
clay

1.0E-08

1.0E-09
1.0E-09 1.0E-08 1.0E-07 1.0E-06 1.0E-05 1.0E-04
laboratory coefficient of permeability k l (m/s)

Fig. 5 The relationship between laboratory and in-situ
coefficient of permeability

100
ratio of k (in-situ) to k (laboratory) k i/k llaboratory coefficient values correspond to the effective
overburden pressure from the relationship between the
consolidation pressure P and the coefficient of permeability k(the log P-log k relationship) obtained in the oedometer test.
The ratio of the field coefficient to the laboratory coefficientof permeability (ki/kl) ranged approximately from 3 to 7 forclay but was larger for peat, mostly ranging approximatelyfrom 10 to 100. Kogure [7] pointed out that areas
surrounding large dead plant and shrub remains in peatground likely became to water flow paths. In oedometer tests
for peat, samples are limited in size and need to be preparedwhile avoiding large dead plant remains. For this reason, thecoefficient of permeability obtained from oedometer tests
fails to evaluate macroscopic water flow paths in peat, andmay underestimate the permeability of the entire soil layer.
The relationship between in-situ void ratio and ki/kl isshown in Fig. 6. ki/kl roughly increases with an increase in
in-situ void ratio. Peat with higher values of in-situ void ratio
contains larger amounts of undecomposed large dead plantre-mains. It can be inferred that this tendency is a result of
accumulation containing a large number of macroscopic porewater flow paths.

k i /k l = 2.4 exp(0.2 e 0)10peat
organic clay
clay
1
0

5

10

15

20

in-situ void ratio e 0

Fig. 6 The relationship between in-situ void ratio andratio of in-situ to laboratory coefficient of permeability

-6

-0.75

ki = 5.3×10 t50

45
measuredpore water pressure U (kPa)pressure to dissipate (min.). It is considered possible toestimate the in-situ coefficient of permeability in peat groundbased on this relationship from dissipation testing of porewater pressure and Eq. (3). Tanaka et al. [8] reported that theaccuracy of pore water pressure estimation is low when Uf iscalculated using the hyperbola method under conditions
without pore water pressure measurement over a sufficientperiod of time. Attention must be paid to the problem when
this estimation method is used.

estimated
40
U 50=34.9kPa
35

U f =28.1kPa

30

t 50=70s

25

(3)

1

10

100

1,000

10,000elapsed time t (s)In this study, tests relating to in-situ permeability wereconducted to determine the permeability characteristics ofpeat ground. The results can be summarized as follows:(1) The in-situ coefficient of permeability in peat ground was
one or two orders of magnitude greater than that of clay ororganic clay, ranging from 1.7×10-7 to 1.1×10-5 (m/s).
(2) The laboratory coefficient of permeability in peat ground
ranged from around 1/10 to 1/100 of the in-situ coefficient.
Matter from large plants and other objects must be
avoided when consolidation testing of peat ground isconducted. Accordingly, the laboratory values obtainedin this test are considered to represent underestimations of
the ground’s macroscopic permeability.
(3) The ratio of the in-situ coefficient of permeability ki to thecoefficient of permeability kl (ki/kl) rose with higher
in-situ void ratio, indicating that the influence of thevolume of water passages in peat ground becomessignificant with greater in-situ void ratio.
(4) The in-situ coefficient of permeability decreased linearlywith higher values of t50, which is the result of the porewater pressure dissipation test using the CPTU method.
This relationship can be used to estimate the in-situ
coefficient of permeability in peat ground.
REFERENCES[1] Hayashi, H., Nishikawa, J., Odajima, H., Mitachi, T. andFukuda, F., Deformation Analysis of Peaty Ground with
Cam Clay Model, Proceedings of IS-Hokkaido ’94,
Vol.1, pp.575-581, 1994.
[2] Schofield, A and Wroth, P., Critical State Soil Mechanics,
McGraw-Hill, London, 1968.
[3] Hayashi, H., Mitachi, T., Tanaka, H. and Nishimoto, S.,
Determination Procedure of Soil Parameters for
Elasto-plastic FE Analysis of Peat, Proceedings of 13thof the Asian Regional Conf. on SMGE, Vol.1,
pp.145-148, 2007.
[4] Robertson, P., Sully, J., Woeller, D., Lunne, T., Powell, J.
and Gillespie, D., Estimating coefficient of consolidationfrom piezocone tests, Canadian Geotechnical Journal, 29,
pp.539-550, 1992.
[5] Tanaka, H., Sakakibara, M., Goto, K., Suzuki, K. And
Fukazawa, T., Properties of Japanese normallyconsolidated marine clays obtained from static piezocone

775

Fig. 7 The typical time history of measured pore waterpressure from CPTU dissipation test
(Riyamunai site at depths of 3.0 m)

1.E-04in-situ coefficient of permeabilityk i (m/s)

6. CONCLUSION

1.E-05

1.E-06

1.E-07k i= 5.3x10-6 t 50-0.75
1.E-08
0.1

1

10

100

1000

t 50 (min)Fig. 8 The relationship between t50 and in-situ
coefficient of permeability
penetration test, Report of The Port and Harbour
Research Institute, 31(4), pp.62-92, 1992. (in Japanese).
[6] Ohira, Y., Peat - engineering properties, investigationand design -, Sekohgizyutsu, 4(6), pp.146-148, 1971. (inJapanese)[7] Kogure, K., Geoengineering of highly organic soils,
Toyo-shoten co. ltd., Tokyo, pp.56-68, 1995. (inJapanese)

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Extraction of temporal and spatial properties on habitat of snail by means ofstatistical approach
Masaaki KONDO and Takamitsu KAJISA
Graduate School of Bioresources, Mie University, Japan
ABSTRACT
A lot of fireflies, which are seen in early summer night, feed on snails. In recent years, public works are increasingly
carried out with sufficient attention given to biology in the way to low-impact. There are water channel suitable for
firefly as ecological zone, which are developed as a way of attention to biological environment. The notion of attention
to biological environment is important in not only city life but rural human life space. National Monitoring ofBiological Indices in Rivers has been conducted for many years in Japan. The data of biology and other indices havebeen accumulated, and it is necessary to analyze the observations. Snails living on the bottom of rivers cannot help but
suffer from variation of discharge such as flood, and decline of water quality at times. These are many biological studies
on snail, but it seems to be less study in terms of hydrological study and water quality. In this study area properties andtemporal properties on snails were obtained by analyzing data set on stream in Mie Prefecture using quantification
theory. Properties of habitat were investigated on Oomatagawa River in the south of Mie Prefecture.
Analysis of data showed useful results on temporal properties and areal properties of snail habitat. It is noted thatquantification theory resulted that snail adjust to the environment of the area such as river width and bed material inriver, which are important parameters. It is expected that potentiality of habitat on snail are evaluated by SOMs.
Keywords: snail, hydraulic, water quality, soil, quantification theory, self-organizing map
1. INTRODUCTION

2. DATA AND METHOD

It is fairly obvious that precipitation affects benthos in river
such as rivers snails, as well as civil parameters such as flow
discharge and bed materials. These are closely related toeach other.
Biological monitoring methods in Japan have been improved,
and their data has been accumulated for many years [1]. Partof the data is made public on the Internet. It is expected thatmonitoring data will be investigated and applied. Whilepublic work is carried out considering the care of rare species
and their habitat, sometimes the care is not sensitive enoughto the environment. Water channels suitable for fireflies
similar to habitats in natural environments are often
developed to enrich human life quality [2]. It is expected thatmonitoring data will be used to develop management
procedures and protection procedures for habitats.
Mie Prefecture is suitable to investigate the impact of rivers
on benthos as shown in Fig. 1. Because Mie Pref. stretches
far to the north and south, and a local region in Mie Pref. has
one of the heaviest rainfalls. Rainfall can affect the habitat ofbenthos in rivers such as snails. In this study, the relation
between civil parameters, habitat, and their properties isgrasped. First, the time series properties and local area
properties are grasped by quantification theory. Next, theproperties of potential habitats are extracted by
self-organizing map (SOM).

2.1 Data and Quantification Theory
2.1.1 Data of Hydraulics, Water Quality, Soil, andBiology
Data of National Monitoring of Biological Indices in Rivers
are offered by Ministry of the Environment [3] in Japan. It ischaracterized that monitoring is carried out on the basis ofFig.1 Location of (A) Mie Pref., (B) local areas, (C)Oomatagawa River, and observation stations in the river(JDG2000)776

Table 2 Measured item and quantify
Hydraulics

Quantity

Depth

Flow velocity

1

0- 5m

0 - 10 cm

Slow/ 0 - 30 cm/s

<10°C

Clear

2

5 - 10 m

10 - 20 cm

Normal/ 30-60 cm/s

10 - 20°C

Low

Fast
/ >60 cm/s

20 - 30°CHigh

3

10 - 20 m

20 - 30 cm

4

>20 m

>30 cm

Year
2004
2005
2006
2007
2008
2009
Total

Soil
Bed material
Head-sized
gravelFist-sized
gravelSmall graveland sand

Quantity

Biology
Snail

0

None

1

Few

2

Many

Sand and mud

6

Others

2.1.2 Rejection and Translation of Datasets
After rejecting incorrect datasets, the qualitative datasets arequantified for analysis. Datasets lacking in analysis items
and a few datasets measured on rainy days were rejectedfrom analysis. Furthermore, the research manual advised
avoiding swollen streams and rivers [4]. Thus, datasets usedin analysis seem to be the ones for normal river conditions. It
appears that turbidity is composed of less soil particle
components and more organic algae component. Next,
datasets suitable for analysis of hydraulics, water quality,
soil, and river snails were translated into integers on the basis
of Table 2.
2.1.3 Precipitation, BOD, and COD
Using data on precipitation between 2004 and 2009 provided
by the Japan Meteorological Agency [5], the annual amountof precipitation was calculated for five regions in Mie. Mie
Pref. publishes monthly data on biochemical oxygen demand(BOD) and chemical oxygen demand (COD), which arereported on the Yearly Data Book of Water Quality in Public
Waters [6]. Annual average BOD and COD are calculated
for five regions in Mie during 2004 and 2008.

Analysis of local area
Number of observation
station for six year
Area A / Hokusei
93
Area B / Iga
14
Area C / Tyusei
37
Area D / Nansei
36
Area E / Higashi-kisyu
90
Local area

Total

>30°C

Turbidity

Wide particle
size distribution

Table 1 Number of observation station
Number of observation
station in all Mie Pref.
50
50
51
37
40
42
270

Water temperature

5

Fig.2 Location of observation stations between 2004 and2009 for analysis, (A) to (F), for time series analysis; (G),for local area analysis (Numbers showed at each local
area are one of station)

Analysis of time series

Water quality

Width

270

2.1.4 Analysis by Quantification Theory
Mathematical quantification theory class II shows the
explanation of external criterion as qualitative throughexplanatory variables shown as qualitative [7], [8]. Civil
parameters such as hydraulics, water quality, and soil
correspond to explanatory variables, and the snail parameteris equal to the external criterion.
The method of mathematical quantification theory class II
was used for time series analysis and local area analysis.
Concerning time series analysis, datasets in all of Mie Pref.
were subjected to analysis one by one year as shown in Fig. 2(A) - (F). the area and data of Mie Pref. is divided into five
regions: Hokusei, Iga, Tyusei, Nansei, and Higashi-Kisyu as
shown in Fig. 1 (B). Data composited for six years for eacharea were analyzed (Fig. 2 (G)).
2.2 Data on Oomatagawa River and Analysis by
Self-Organizing Map
The studying of potential of habitat focused on only
Oomatagawa River as shown in Fig. 1 (B) and (C). The factthat there were few snails in the rivers of Higashi-Kisyu
indicates that the environment will be sensitive to the effects

reliable procedures and not only organism species in rivers
but also civil parameters are measured. Almost all of themonitoring was carried out in the summer vacation because
many elementary and junior high school students
participated in the monitoring. They identified organism
species with the assistance from experts in biology.
The data on rivers selected in all of Mie Pref. were analyzed
between June and September between 2004 and 2009. Fig. 2
shows the location of observation stations for analysis. Table
1 shows the number of observation stations. A total of 270
stations for 58 rivers were investigated for six years.
The number of stations where one-year analysis was used
was between 37 and 51 stations. The number of station of thelocal area analysis totaled between 14 and 93 stations during
2004 - 2009.
Hydraulics parameters such as river width, depth, and flow
velocity, and water quality parameters such as temperature,
turbidity, and soil parameters of bed material are selected foranalysis from the many items measured in civil parameters.
We analyzed the river snail as benthos.
777

(A)

Few(1)

None(0)Many(2)

100%

50%

0%
2004

2005

2006

2007

2008

2009

None(0)Few(1)

Amount of river snail, %

4.0

2005

3.5

2005

3.0

2007

2007
2008

2.5

Average

2.0
1.5
1.0
0.5
0.0
Area A /HokuseiMany(2)

Area B / Iga

Area C /Tyusei

Local area

Area D /Nansei

(B)

2004
2006
2008

2.5

Average

2.0
1.5
1.0
0.5
0.0

Area E /Higashi-kisyu

Area B / Iga

Area A /Hokusei

Area D /NanseiArea C /Tyusei

Local area

Area E /Higashi-kisyu

Fig.5 (A) Annual-average BOD concentration, (B) annual-average COD
concentration in local area

100%

50%

Table 3 Data list of Oomatagawa River for SOM analysis
0%

Year

Area A /Hokusei

Area B /Iga

Area C /TyuseiArea D /NanseiArea E /Higashikisyu

Local area

2004

Fig.3 (A) Change of amount of river snail
with time in all Mie Pref., (B) amount ofriver snail in local area for six year.

Annual precipitation, mm/yr

2004
2006

3.0

Year

(B)

(A)

3.5

Annual-average COD, mg/l

Annual-average BOD, mg/l

Amount of river snail, %

4.0

2004
2006
2008
Average

4500
4000
3500
3000
2500
2000
1500
1000
500
0

2005
2007
2009

2005

2006

2007

2008

Area A /Hokusei

Area B / Iga

Area C /Tyusei

Local area

Area D /NanseiArea E /Higashikisyu

Fig.4 Annual precipitation in local area

2009

Observation station

Asuka elementary school
Tyobokujyou lumberyard
Shikinosato
Genbokuichiba market
Sawataribashi bridge
Nishin elementary school
Itsusato elementary school
Itsusato elementary school
Tyobokujyou
Asuka elementary school
Shikinosato
Genbokuichiba
Sawataribashi Bridge
Nishin elementary school
Itsusato elementary school
Itsusato elementary school
Asuka elementary school
Tyobokujyou
Shikinosato
Genbokuichiba
Nishin elementary school
Sawataribashi Bridge
Itsusato elementary school
Itsusato elementary school
Itsusato elementary school
Itsusato elementary school
Nishin elementary school
Sawataribashi Bridge
Asuka elementary school
Tyobokujyou
Marubuchi
Itsusato elementary school
Itsusato elementary school
Asuka elementary school
Marubuchi
Noguchi
Itsusato elementary school
Itsusato elementary school
Marubuchi
NoguchiTyobokujyou
Asuka elementary school
Ikedataira

StationMark atnumber at
Fig.6
Fig.1
St.1
St.2
St.3
St.4
St.5
St.6
St.7
St.7
St.2
St.1
St.3
St.4
St.5
St.6
St.7
St.7
St.1
St.2
St.3
St.4
St.6
St.5
St.7
St.7
St.7
St.7
St.6
St.5
St.1
St.2
St.8
St.7
St.7
St.1
St.8
St.9
St.7
St.7
St.8
St.9
St.2
St.1
St.10

★飛04
貯04
★四04
原04
佐04
☆日04
五04
五04
貯05
☆飛05
☆四05
原05
佐05
☆日05
五05
五05
☆飛06
貯06
★四06
原06
☆日06
佐06
五06
五06
五06
五06
★日07
佐07
☆飛07
貯07
☆ま07
五07
五07
★飛08
★ま08
☆野08
五08
五09
☆ま09
★野09
貯09
☆飛09
池09

1
WT

2
3
Width Depth

°C

m

cm

22
20
23.5
24
23.9
22
24.7
26.7
18
19
21
24
21
19
23.5
23.5
20.5
19
21.5
23
20.5
22
22.5
25
23.5
25.5
19
18.8
20.5
19
22.5
24.8
24.5
18.2
21.8
19
27
19.2
18
19
16.8
18.7
21

10
10
8
10
20
13
30
30
8
10
10
15
10
10
15
15
10
10
10
10
18
10
15
15
10
8
20
15
10
10
15
15
15
10
10
20
10
10
10
20
10
10
20

30
25
30
30
20
20
40
30
15
15
20
15
10
15
20
20
20
20
20
20
25
10
40
20
15
15
20
20
25
25
30
40
20
20
25
25
35
40
25
25
25
20
30

4

5
6
Flow velocity

Slow Normal Fast
0
0
0
1
1
0
0
1
0
0
0
1
0
0
0
1
0
0
0
1
0
0
1
1
1
1
0
0
0
0
0
1
1
0
1
0
1
1
1
0
0
0
0

1
1
1
0
0
0
1
0
1
0
0
0
1
1
1
0
0
1
0
0
0
1
0
0
0
0
0
1
0
1
1
0
0
1
0
0
0
0
0
1
0
1
1

0
0
0
0
0
1
0
0
0
1
1
0
0
0
0
0
1
0
1
0
1
0
0
0
0
0
1
0
1
0
0
0
0
0
0
1
0
0
0
0
1
0
0

7
Headsized
gravel
1
0
0
1
0
1
0
0
0
0
0
0
0
0
1
0
1
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
1
1
0
0
0
0
0

8

9

10
Bed material

11

12

Small
FistSand
Wide particle
gravelsized
Othersand mud size distribution
gravel and sand
0
0
1
0
1
0
1
1
0
0
0
0
0
1
0
0
0
1
0
0
0
0
1
1
0
0
0
0
0
0
0
0
0
0
0
1
0
0
0
0
0
0
1

0
1
0
0
0
0
0
0
0
0
0
0
0
0
0
1
0
0
1
1
1
0
0
0
1
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

0
0
0
0
0
0
0
0
1
1
1
1
1
0
0
0
0
0
0
0
0
1
0
0
0
1
1
1
1
1
1
1
1
1
1
0
0
0
1
1
1
1
0

0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

13

14
15
Turbidity

16

17
Snail

18

Clear

Low

High

None

Few

Many

1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
0
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1

0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
1
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

0
1
0
1
1
1
1
1
1
1
1
1
1
1
1
1
1
1
0
1
1
1
1
1
1
1
0
1
1
1
1
1
1
0
0
1
1
1
1
0
1
1
1

1
0
1
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
1
0
0
0
0
0
0
0
1
0
0
0
0
0
0
1
1
0
0
0
0
1
0
0
0

0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0
0

Note; WT=Water temperatureon snails. A lot of monitoring points were set up and havecontinued to investigate the situation in Oomatagawa River.
Oomatagawa River is suitable for investigating the existenceof habitat.
It is noted that only Oomatagawa River was focused on thancomparing various rivers because of avoiding the particulareffect of pesticide, releasing snail in river, and heavy metals[9]. Furthermore, it is difficult to analyze rivers of manysnails because of keeping the same condition.

Analysis data are normalized in most of the SOM study. In
this study, data are normalized so that the mean is equal tozero and dispersion is equal to one.
2.2.2 Analysis by Self-Organizing Map
A self-organizing map (SOM, also called Kohonen’s map)
was used as an analysis method [10], [11]. The SOM
produced a two-dimension map from multidimensional
elements on monitoring data, and it visualized classification.
Kohonen et al. (1995) of Helsinki University of Technology
developed the software of SOM_PAK called Basic SOM andpublish it on their website [12]. The 15-dimension data of
Oomatagawa River on factors such as hydraulics, water
quality, and soil property were applied to the analysis. The
class of observation station is visualized to consider the
results.
Values of learning parameters for SOM were set as follows:
map size is 15 in the x-direction and 10 in the y-direction,
limit on time iteration is 10,000, learning restraint is 0.05,
and neighborhood function is 21.

2.2.1 Data on Oomatagawa River
Table 3 lists the monitoring data of observation stations for
Oomatagawa River shown in Fig. 1 (C). The data are
selected between June and September between 2004 and2009 to grasp the recent trend. Parameters, such as riverwidth, depth, and water temperature in Table 3 arecontinuous values, the others are flag values. The parameterof river snails is excluded from analysis. The analysis datahas 15 dimensions. The values of BOD and COD are notmeasured in individual observation stations.
778

Table 4 Result of time series analysis ((A), Item; (B), detail; (C), total evaluation)
Ranking and score (SA ) of items

(A)

No.2

No.3

No.4

No.5

No.6

6 pt

5 pt

4 pt

3 pt

2 pt

1 pt

2004

Bed

Turbidity

Depth

Flow

Width

WT

2005

Bed

Depth

Turbidity

Flow

Width

2006

Bed

Turbidity

Width

Flow

Depth

2007

Bed

Turbidity

Width

Depth

Flow

2008

Width

Bed

Flow

Turbidity

Depth

2009

Width

Bed

Turbidity

Depth

Ranking of details (DB )

(B)

No.1

Year

Flow

No.1

No.2

No.3

No.4

No.5

No.6

2004

Wide particle
size distribution

Cloudy20-30cm

Normal0-5m20-30°C

WT

2005

Sand and mud

>30cm

Dirty

Normal0-5m20-30°C

WT

2006

Others

Dirty

0-5m

Fast

WT

2007

Sand and mud

Clear

0-5m

10-20cm

Normal20-30°C

WT

2008

0-5m

Sand and mud

Fast

Cloudy>30cm

10-20°C

0-5m

Wide particle
size distribution

Clear

0-10cmSlow

20-30°C

Year

WT

2009

20-30cm 10-20°C

Note; Bed=bed material; WT=water temperature; Flow=flow velocity(C)

Details (Xi ) and each score (SC )

Item
Width

Depth

Flow

WT

Turbidity

Bedmeets the level of drinking water needing normal
purification. Water in all of Mie Pref. is recognized as clear;in particular, water from Higashi-Kisyu, showed in Fig. 1(B) as Area E, is clear. Therefore, the organic matter onwhich snails feed is less in Higashi-Kisyu.

0-5m

5-10m

10-20m

>20m

24pt

0pt

0pt

0pt

0-10cm

10-20cm 20-30cm

3.2 Properties of Time Series Using Quantification
TheoryTable 4 indicates the results of the analysis. Bed material,
river width, and turbidity seem to contribute highly to snail
habitat as shown in Table 4(A). It is difficult to understandthe properties without including the data shown in Table
4(B). To provide the scores depending on the order, Table
4(C) was rearranged using the following equation (1):

>30cm

3pt

3pt

6pt

7pt

Slow

Normal

Fast

2pt

8pt

7pt

<10°C

10-20°C

20-30°C

>30°C

0pt

2pt

4pt

0pt

Clear

Low

High

9pt

8pt

9pt

Wide
Head-sized Fist-sized Small gravel Sand andparticle sizeand sandgravelgravelmud
distribution

0pt

0pt

0pt

17pt

11pt

SC ( X i ) = ∑ {SAyr × Y ( X i − DB )}

(1)

yr

Others6pt

where SC is a score of Xi, Xi is a detail in Table 4(C), DB is adetail in Table 4(B), yr is year, SA is the score in Table 4(A),and Y is the step function; if Xi = DB, Y = 1, and if Xi is notequal to DB, Y = 0.
Table 4(C) shows that the properties of the time series
resulting from aggregating the contents shown in both Table
4 (A) and (B) have greater accuracy. 0-5 m of river width (24points), sand and mud bed material (17 points), clear or highturbidity (both 9 points), and other items such as flow
velocity, depth, and water temperature are presented indescending order of points.
First, the top three of all items: river width, bed material, andturbidity are explained as follows. The relatively narrow 0-5
m river width corresponds to a flume or small stream inJapan. Snails prefer narrow rivers into wide rivers(Murakami (2007) [14]). The analysis results agree with his
description.
Sand and mud bed material showed high point. Furujo andTomiyama (2000) [15] indicated that snails prefer gravel,sand, and mud. Murakami (2007) [16] also indeicated thatthat snails in upstream sites feed on periphyton on gravel andwithered leaves. It is considered that since the light material
such as sand, mud, and withered leaves are easy to sedimentin the condition of low velocity, snails prefer the
sedimentary surrounding.

Note; Bed=bed material; WT=water temperature; Flow=flow velocity

3. RESULT AND DISCUSSION
3.1 Characteristics for Analysis Data Based on Statistics
The modal class on civil parameters is: 5-10 m of river width,20-30 m of depth, normal flow velocity, fist-sized gravel bed
material. Data selected during the summer observation
caused the water temperature to be 20-30°C at most, andthere was no data for below 10°C. Clear turbidity was mostcommon, low turbidity was present in 52 out of 270, andhigh turbidity, which was rare, was present in three out of
270.
Fig. 3 indicates the amount of river snails for the time series
and each local area. River snails were observed in many sites
in 2008 as shown in Fig. 3 (A). Iga and Tyusei, Hokusei andNansei, and Higashi-Kisyu were ranked in descending orderof snails in each area. There were fewer snails in
Higashi-Kisyu/Area E than in other areas.
Fig. 4 shows the amount of rainfall in each local area. In
accordance
with
the
local
area,
rainfallin
Higashi-Kisyu/Area E, where it naturally rains a lot, ismarkedly high.
Fig. 5 shows annual average BOD and COD in rivers. BOD
and COD are low in all of Mie pref. In accordance with theenvironmental quality standard in Japan [13], river water779

Table 5 Result of local area analysis ((A), Item; (B), detail)Ranking and score of items(A)
Local area

Ranking of details

(B)

No.1

No.2

No.3

No.4

No.5

No.6

Area A /Hokusei

Bed

Turbidity

Width

Flow

Depth

WT

Area B /
Iga

Bed

Flow

Width

Depth

WT

Area C /Tyusei

Bed

Width

Depth

Flow

Area D /Nansei

Depth

Bed

Width

Area E /Higashi-kisyu

Turbidity

Depth

Bed

Local area

No.1

No.2

No.3

No.4

No.5

Area A /Hokusei

Sand and mud

Low

0-5m

Slow

20-30cm 20-30°C

Turbidity

Area B /Iga

Head-sized
gravelFast

0-5m

0-10cm

10-20°C

Low

Turbidity

WT

Area C /TyuseiWide particle
size distribution

5-10m

10-20cm

NormalHigh

20-30°C

Flow

Turbidity

WT

Area D /Nansei>30cm

Others

5-10m

Slow

Low

20-30°C

WT

Width

Flow

Area E /Higashi-kisyu

Low0-10cmFist-sized
10-20°Cgravel>20m

NormalNo.6

Note; Bed=bed material; WT=water temperature; Flow=flow velocity

The effect of turbidity as feed will be an important factor forsnails because non-clear turbidity, summing up eight pointsof low and nine points of high turbidity, is more than the nine
points of clear turbidity. Non-clear turbidity affects the
habitat of snails.
Next, Flow velocity, depth, and water temperature in the low
ranks are explained. Snail on river beds can be flushed away
at flow velocities of more than one m/s [17]. Snails areobserved in riffles with high velocities [15]. It does not meanthat high velocity is a bad condition. Thus, the fact that theflow condition is uncertain leads to a low position in Table 4.
Snails are fed on by enemies such as fireflies, river crabs,
rats [18]. Deep water may protect snails from these enemies.
However, snails are observed regardless of the depth of theriver [15]. It is considered that the habitat condition on depth
is ambiguous.
The reason why data of water temperature was restricted insummer caused to be low points in Table 4.

Fig.6 Self-organizing map on observation stations in
Oomatagawa River (Marks in the figure are showed in
Fig.1 and Table 3)
considered as a feed index.
3.4 Results using SOM
Fig. 6 indicates the SOM on Oomatagawa River. A black starmeans that the snail was captured, a white star means thesnail was not captured but was captured in another year, nomark means the snail was not captured. Not-captured sites
like Itsusato are located on the left side of the map. Most of
the black stars, all of the white stars, and a few of thelocations with no marks such as Sawataribashi Bridge andTyobokujyou lumberyard are located on the right side. The
right side is under the mixed conditions.
The results of analysis are interpreted as follows; it is almost
impossible to observe snails on the site of left side because
almost all were not captured. However, it is highly possible
to observe snails on site of the right side because of themixing conditions.
Snails have not been observed at the Itsusato site. It will be
hard to observe snails there. The Itsusato site is located on
nearby reservoir and downstream with wide river width (Fig.
1(C)). Thus, it is considered that sand accumulated aroundthe Itsusato site damages periphyton on the gravel surface
and no snails are observed.

3.3 Properties of Local Area Using Quantification
TheoryTable 5 indicates the results of local area analysis. As shownin Table 5(A) and (B), the analyzed order of local area
resulted in difference from each other. Area A / Hokusei hassand and mud bed material, low turbidity, and 0-5 m riverwidth in sequence. Area B / Iga has head-sized gravel bed
material, fast flow velocity, and 0-5 m river width. Area C /Tyusei has wide particle size distribution bed material, 5-10
m river width, and 10-20 cm depth. Area D / Nansei has over
30 cm depth, other bed material, and 5-10 m river width.
Area E / Higashi-Kisyu has low turbidity, 0-10 cm depth,
and fist-sized gravel bed material.
Thus, the property of a local area is similar to the property oftime series regarding of an importance of river width, bed
material, and turbidity. The property of Higashi-Kisyu isdifferent from the others; turbidity resulted in highcontribution toward the habitat of snails.
There are few snails in Higashi-Kisyu as described in section
3.1. Snails in clear or pure rivers like the upper portion of ariver feed on periphyton on gravel or spoiled leaves [16]. A
lot of rainfall causes damage to the surface of gravel by theirrolling, and it is hard for periphyton to grow on gravel. This
may result in few snails and in less BOD and COD

4. CONCLUSION
In this study, it was investigated that the civil parameters780

exerted on habitat of river snail and the possibility of habitat.
Data of civil parameter on rivers and snail concerning Mie
Pref. in Japan were subject to the analysis by the method ofboth quantification theory and self-organizing map.
1) Relationship between habitat of snail and quantification
theory was investigated. Compared with references
published by the experts and their experience, river width,
bed material, and turbidity made a large contribution tohabitat.
2) Possibility of habitat for river snail is showed on a SOM.
It is possible to distinguish between easy habitat and difficult
one to live. It suggested that the rolling of gravel andsedimentation cause bed variable to make hard habitat forcontinuously-growing snail.

2004, p.35.

REFERENCE[1]

Tanida K, "Index biology on river environment", Tokyo, Hokuryukan
Publishing, 2010, pp.142-151.
[2] Sekine M, Goto M, Ito N, Tanaka K, Kanao M, and Inoue T,
Construction of a firefly stream by using a physical habitat evaluation
method, Ecol. Civil Eng. 10(2), pp.103-116, 2007.
[3] Ministry of the Environment (2004-2009), "Data of National
Monitoringof
Biological
IndicesinRivers",<https://www2.env.go.jp/water-pub/mizu-site/mizu/suisei/>, browsed
on April 1st, 2012.
[4] Ministry of the Environment and Ministry of Land, Infrastructure,
Transport and Tourism, "Kawa no ikimono wo sirabeyou (researchmanual)", Japan Society on Water Environment, 2000.
[5] Japan Meteorological Agency (2004-2009), "Table of Hourly Weather
Observations", < http://www.jma.go.jp/en/amedas_h/map38.html>,
browsed on April 1st, 2012.
[6] Mie Pref., "Yearly Data book of Water Quality in Public Waters", Mie
Pref., 2004-2009.
[7] Hayashi C., "On the quantification of qualitative data from themathematico-statistical point of view -An approach for applying thismethod to the parole prediction-", Ann. Inst. Statistical Math., 2,
pp.35-47, 1950.
[8] Hayashi C., "On the prediction of phenomena from qualitative data
and the quantification of qualitative data from themathematico-statistical point of view", Ann. Inst. Statistical Math., 3,
pp.69-98, 1951.
[9] Iskizaki S and Hamada H, “Effects of Heavy Metals on the Freshwater
Snail, Semisulcospira bensoni, in a Closed Mining Area”, Jpn. J.
Limnol., 48(2), pp.91-98, 1987.
[10] Kohonen T, "Self-Organizing Maps", Tokyo, Springer, 2010.
[11] Ookita M, Tokutaka H, Fujimura K, and Gonda E, "Self-Organizing
Maps and its Software", Tokyo, Springer, 2008.
[12] Kohonen T, Hynninen J, Kangas J, and Laaksonen J, "SOM_PAK-Self-Organizing Map Program Package", Helsinki University ofTechnology, 1995.
[13] Announcement by Ministry of the Environment (1971),"Environmental standards concerning water pollution -appended table2-", <http://www.env.go.jp/kijun/mizu.html>, browsed on April 1st,
2012.
[14] Murakami M, "How to breed and to photograph firefly in freshwater",Powersha, 2007, p.87.
[15] Furujo Y and Tomiyama K, "Distribution and Microhabitat of
Coexisting Two Freshwater Snail Species, Semisulcospira libertina(Gould) (Prosobranchia; Pleuroceridae) and Clithon retropictus(Martens)(Prosobranchia : Neritidae)", Venus : the Japanese j. ofmalacology 59(3), pp.245-260, 2000.
[16] Murakami M, "How to breed and to photograph firefly in freshwater",Powersha, 2007, p.4.
[17] Mori S, "Some ecological notes on the fresh water snails - second -",Venus, 6(1), pp.15-21, 1936.
[18] Tokyo Fireflies Ecology Institute, "Hotaru hyakka", Tokyo, Maruzen,

781

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The use of Nano Zero Valent Iron to remediation of contaminated soil andgroundwater
Taghizadeh Maryam 1, Yousefi Kebria Daryoush 2 and Gholamreza Darvishi 3
Babol University of Technology, Iran

ABSTRACT
Environmental contaminants have been known to be present in many hazardous waste sites, which madean enormous impact on the quality of groundwater, soil and associated ecosystems. Remediation of thesecontaminated sites is an important challenge facing the scientific and technical community.
Nanotechnology is the use of manipulating matter on an atomic at least one dimension sized from 1 to
100 nanometers and molecular scale. Nano Zero Valent Iron (nZVI) is emerging as new option for thetreatment of contaminated soil and groundwater. Due to their small size, the particles are more reactive
than granular iron which is conventionally applied in reactive barriers and can be used for in situ
treatment. nZVI effectively reduces chlorinated organic contaminants (e.g. PCB, TCE, PCE, TCA,
pesticides, solvents) and also inorganic anions (perchlorate). This present research gives an overview
over the characteristics and application of nano zero valent iron (nZVI) in general and summarizes of theexperiences on groundwater remediation with other scientists.

Keywords: Remediation – Nanotechnology - groundwater - Nano Zero Valent Iron (nZVI)
1. INTRODUCTION
Maintaining and restoring the quality of air, water and soil is
one of the great challenges of our time. Most countries faceserious environmental problems, such as the availability ofdrinking water, the treatment of waste and wastewater, air
pollution and the contamination of soil and groundwater.
Enormous effort has been made to find efficient and effective
ways to remediate petroleum contamination in soil andgroundwater [1].
Petroleum compounds and its derivatives are cause
contamination groundwater and soil, however petroleum
materials penetrated to more depth of soil, decontamination
will be more difficult. Techniques that are used for cleanup of
soil and ground water include the use of water for to remove
pollutants, chemical solvent, incineration and help toorganism to break down the pollutant molecules.
Thousands of sites exist in the United States and worldwide
where subsurface soils and groundwater have been
contaminated with a wide range of toxic organic compounds
and heavy metals.
In particular, halogenated organic contaminants, such asPentachlorophenol, Trichloroethylene, Trichloroethane,
Dinitrotoulene, and Trinitrotoluene, are present at several
sites, and these contaminants are listed as priority pollutantsby the United States Environmental Protection Agency(USEPA) due to their toxicity and carcinogenicity.
These contaminants are persistent in the environment slowly
by natural processes [2].

The conventional methods of treating these contaminants
include soil washing/ flushing, thermal desorption,
vitrification, and bioremediation; however, these methods are
relatively expensive, slow, or limited by the production of
secondary waste streams that require subsequent disposal ortreatment[3].
Nanoscale iron particles represent a new generation ofenvironmental remediation technologies that could provide
cost-effective solutions to some of the most challenging
environmental cleanup problems. Nanoparticles show ahigher catalytic activity because of their small size (10-100nm) and their large specifiic area.
Nanotechnology is a branch of applied sciences that can be
defined as particles of size ranging from one to hundred
nanometers (nm) in any dimension. The size of nano particles
is of several times smaller than even the red blood cells.
Nanotechnology is very diverse, ranging from extensions ofconventional device physics to completely new approaches
based upon molecular self-assembly, from developing newmaterials with dimensions on the nanoscale to direct controlof matter on the atomic scale [4].
Andreta is showed that nanotechnology is the actions andreactions that occur at the atomic level, a more accurateinterpretation nanotechnology is new revolution for all
science in the future. This technology can improve methods
assessment , management and reduce risks for theenvironment and will provide opportunities for new products[5].
2. NANOTECHNOLOGY APPLICATION IN
ENVIRONMENTAL
Nanotechnology has provided the possibility of producing
product with high quality, low cost and high speed. One of theimportant areas of application of nanotechnology in the782

environmental is removal or cleanup of chemical
contaminants such as chemical pesticides or convert them tosubstances that have less risk of toxicity. For example, can beused of nano zero valent iron for the cleanup of soil andgroundwater for a variety of chemical contaminants. These
nanoparticles as a catalyst was caused accelerate theoxidation process and cause the Organic pollutants suchas chemical pesticides are broken down into smaller carbon
compounds that have less toxicity. The same projects about
removal of microbial and chemical contaminants of water,
soil and air is in progress in anotechnologies centers at
various countries around the world, that represent thecapabilities of nanotechnology in this field.
2.1 ENVIRONMENTAL CLEANUP OF
HYDROCARBON CONTAMINANTS
There are different methods for improvement and cleanup of
soil and groundwater, for example in pump and treat method,which is one of the conventional methods, groundwater
pumped and after filtration and removal of contaminants,
groundwater aquifers are returned again, this method is very
costly and take a long time.
Iron was first recognized and patented in 1972 as achlorinated pesticide degrader. In 1981, Sweeny utilized ironpowders to degrade various hydrocarbons, such astrichloroethylene [6]. Additional suggestions for using zero
valent iron to degrade trichloroethylene and trichloroethane
were made in the late 1980,s by Senzaki (1988) [7]. In 1993 apatent was lodged by the University of Waterloo for using
zero valent iron for treating contaminated groundwaterin-situ, demonstrating the identification of zero valent iron asa remediation constituent.
Nanoscale zero valent iron (NZVI) is more effective atreaching deep zones of contamination, and is more effectiveat contaminant degradation than iron of larger size. (NZVI)
can induce greater rates of reaction because of its greater
specific surface area, which allows a greater exposure of theiron particle to the contaminant per unit weight of iron thanother larger particles. Additionally, as particle size decreases
and tend towards 10 nm, thermodynamic properties, such as
work-function and free energy being to alter and can increase
reactivity. Gavaskar et al has found that NZVI is significantly
more reactive than granular iron, and states that it canremediate a plume in a much shorter time scale [8]. Henry et
al states that nanoscale zero iron has a superior pore
penetration ability when compared to larger particulate zero
valent iron [9].
Zero valent iron has been shown to react and degrade many
types of chemicals[8], including halogenated aliphatics,
polyhalogenated aromatics and nitrates and trichloroethene[10],[9].
The standard half reaction for zero valent iron reacting toyield a ferrous cation and 2 electrons is:(1)This reaction has a standard reduction potential of +0.44 V[11].
Alkyl halides have a typical half reaction as such, where RX
indicates a halogenated hydrocarbon, and X- represents a
halogen anion:(2)
These types of half reactions have reduction potentials
ranging from +0.5 V to +1.5V at pH 7 (Matheson 1994), thevariation is attributed to the wide range of alkyl halides that
this reaction applies to.
When combined, these two half reactions yield athermodynamically spontaneous reaction:(3)
This constitutes the most basic mechanism for halogenatedhydrocarbon degradation by zero valent iron, yielding aferrous cation, an aliphatic hydrocarbon and a halogen anion(Matheson 1994).
A second mechanism for degradation of halogenatedhydrocarbon by zero valent iron is the oxidation of zero
valent iron to a ferrous cation by water (Matheson 1994). The
ferrous ion then further oxidizes to a ferric cation by thefollowing half equation:(4)
This oxidation reaction can be coupled with the reduction half
equation to reduce the alkyl halide shown above to yield:(5)This is a second mechanism for the degradation of an alkyl
halide (Matheson 1994) by zero valent iron.
Matheson (1994) describes a third mechanism, which
involves the zero valent iron reacting with water to yield theferrous cation, the hydroxyl anion and hydrogen gas (H 2 ). It
is combination of the two following half reaction:
P

P

R

R

(6)
(7)
(8)
The H 2 gas generated can then continue on to react with analkyl halide in a reaction known as addition reaction to yield adehalogenated aliphatic, a halogen anion and a proton in thefollowing manner:
R

R

(9)
RX+ H 2 → RX + XIt is important to note that the H 2 can only react with the alkyl
halide if a suitable catalyst is present.
R

R

P

R

R

3. ADVANTAGES OF ZERO VALENT IRON
NANOPARTICLESUsing zero valent iron has the following advantages as a
remediation technology [10]:I. It is relatively inexpensive.
II. It is non-toxic.
III. It degrades certain chemical faster than other techniquesof remediation, such as biotic remediation
IV. It has high energy effectiveness.

783

A widely used technique to treat contaminated groundwaterin-situ is by utilizing Passive Reactive Barriers (PRB). The
location of the PRB must first be ascertained, then theexisting soil must be excavated and the void space filled withthe reactive medium with relatively high hydraulicconductivity.
Some of the initial field testing of PRBs was done using zero
valent elemental iron filings [12],[13]. Incorrectunderstanding of the frequently complex hydrogeology ofvarious contamination sites can lead to incorrect barrier wall
placement, which can leave contaminated zones outside of
the barriers untreated.
Passive Reactive Barrier have proven to be effective at
treating a great number of contaminated groundwater plumes;
they have certain limitations [14]:I. They only target contaminant plume, and not the sourceof contamination. They therefore have to wait for thecontaminant to be leached into or advected with thegroundwater before treatment can be initiated.
II. They are unfeasible solutions in certain situations ofcomplex hydrogeological conditions, such as fractured
rock.
III. Most Passive Reactive Barriers have been installed to adepth of approximately 15 meters, although there havebeen instances of depths up to 35 meters[9]. They cannot
penetrate deep into the soil, rendering them wholly
ineffective with deep plumes, as they cannot reach thetarget zone.
A comprehensive understanding of the hydrogeologicalconditions at the contamination site is required for thistechnology to work, as the positioning of PRB is of utmost
importance.
4. SOIL AND GROUNDWATER REMEDIATIONWITH NANOPARTICLES
Environmental remediation methods can be classified asadsorptive and reactive and as in situ or ex situ [15],[16]. The
use of nanomaterial in all these scenarios has been
investigation. In soil and groundwater remediation, for in situ
treatment, it is necessary to create either an in situ reactive
zone with relatively immobile nanoparticles or a reactive
nanoparticles plume that migrates to contaminated zones.
There are two ways to use nZVI in groundwater and soil
remediation[15],[17]:• nZVI is injected to form a reactive barrier of ironparticles.
• nZVI is injected in surface- modified form to establish
a plume of reactive iron, which destroys any organic
contaminants within the aqueous phase
Several studies have shown that nZVI as a reactive barrier isvery effective in the reductive degradation of halogenatedsolvents, such as chlorinated methane, brominated methane,trihalomethane, chlorinated ethene, chlorinated benzenes andother polychlorinated hydrocarbons, in groundwater [18],[19]. nZVI has also been shown to be effective against
pesticides and dyes [18]. Efficient removal by nZVI ofpolycyclic aromatic hydrocarbons (PAHs) adsorbed to soils
has been reporter at room temperature [20],[21], while under
the same conditions only 38% of the polychlorinatedbiphenyls (PCBs) were destroyed because of the very strong
sorption of PCBs to the soil matrix [22].
Krishna investigated electrokinetic delivery of nanoscale ironparticles for remediation of pentachlorophenol in clayey soil.
The results showed that 80 to 98% PCP was removed fromthe soil within an hour, but PCP reduction was increased from50 to 78% at 1h to 40 to 90% at 24h reaction time for different
NIP concentrations. There was no significant effect of NIPs
concentration on the PCP removal, but the amount of PCP
reduction increased with increase concentration of NIP.
According to Tratnyek and Johnson (2006), nZVI used inreal- world groundwater remediation has a particle size larger
than 100 nm and is thus strictly speaking outside the standard
definition of NP size. These authors also state that themobility of nZVI will be less than a few meters under almost
all relevant conditions as nZVI tends o aggregate, producing
clusters that may approach several micrometers in size andthus be easily removed from the pore water[15]. Companies
are therefore functionalizing nZVI particles to stop them fromaggregating, for example, with surfactants or polymers. Other
approaches combine the nZVI with carbon plateles or embed
the nZVI in oil droplets to facilitate particle delivery into thecontaminated area [19]. In the United States, it is common tocombine the nZVI with other metals, such as palladium, toincrease the reactivity. In Europe such bimetallic particles are
not used due to their possible toxicity and the limited
additional benefit [19].
Tratnyek and Johnson (2006) state that high reactivity tends
to correlate with low selectivity. For this reason, remediation
with nZVI may be inefficient because nZVI particles may
react with non-target substances, including dissolved oxygen,sulphate, nitrate and water. This also implies that nZVI will
have a limited lifetime in porous media and reinjection ofnZVI may be necessary, which makes the treatment morecostly[15].
5. CONCLUSION
Emergence of nanotechnology, particularly synthesis ofnanoscale iron particles (NIP), has provided opportunities todevelop innovative site remediation technologies. The small
particles have great sorption capacity. However, technicalchallenges, such as the delivery of the particles to the target
area, have to be solved.
The reaction pathways of NIP with target halogenatedorganic contaminants are similar to that of zero-valent iron
(iron filings) commonly used in a permeable reactive barrier
technology. However, due to their infinitesimally small size,
NIP can be highly reactive due to their high surface to volume
ratio and greater number of reactive sites and higher intrinsic
reactivity on reactive sites. In addition, NIP can be injected
directly into the contaminated zones, making the in-situ
remediation faster and effective. Due to their small size, theparticles are very reactive (more reactive than granular iron784

which is conventionally applied in reactive barriers) and canbe used for in situ treatment. NZVI effectively reduces
chlorinated organic contaminants (e.g. PCB, TCE, PCE,
TCA, pesticides, solvents) and also inorganic anions(perchlorate). It can even be used to recover/remove
dissolved metals from solution (e.g. Cr (VI), U (VI)). Also
nanoscale zero valent iron is more effective at reaching deepzones of contamination, and is more effective at contaminantdegradation than iron of larger size.

[15]

6. REFERENCES

[19]

[1]
[2]

[3]

[4]
[5]
[6]
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[9]
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Rickerby, D. and M. Morrison, Report from theWorkshop on Nanotechnologies for Environmental
Remediation, JRC Ispra. 2007.
Vogel, T.M., Criddle, C.S. and McCarty P.L.,
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HalogenatedAliphatic
Compounds Environmental science & technology,
1987. 21(8): p. 722-736.
Sharma H.D., R., K.R., Geoenvironmental
Engineering: Site Remediation, Waste Containment,
and Emerginig Waste Management Technologies.
John Wiley, Hoboken, NJ, 2004.
Breytenbach, J.H., The Metris of vaccinationinternational poultry production. Savian Influenza
Control, 2005. 13(4): p. 15-17.
Andreta, E., Nanosciences and Nanotechnologies:
What Future for Research. rfuture Conference andExpo, Chiba- shi, Chiba, Tokyo, Japan, 2003.
Sweeny, K. and J. Fischer, Reductive degradation of
halogenated pesticides. 1972.
Senzaki, T., " Removal of Chlorinated Organic
Compounds by Iron and Manganese Powders in
BufferedWater
andLandfill
Leachate"
Chemosphere, 1988. 29: p. 1743-1753.
Gavaskar, A., L. Tatar, and W. Condit, Cost andperformance report nanoscale zero-valent irontechnologies for source remediation. 2005, DTIC
Document.
Henry, S.M. and S.D. Warner, Chlorinated solventand dnapl remediation innovative strategies for
subsurface cleanup. Recherche, 2003. 67: p. 02.
Zawaideh, L., C. Chew, and T. Zhang. Remediationof nitrate-contaminated water and soil by
Fe0-promoted processes. 1997.
Atkins, P., Physical Chemistry. 6th. 1998, Oxford
University Press, Oxford, Melbourne, Tokyo.
Gillham, R.W., S. O'Hannesin, and W. Orth. Metalenhanced abiotic degradation of halogenatedaliphatics: Laboratory tests and field trials. 1993.
Gillham, R.W. and S.F. O'Hannesin, Enhanced
degradation of halogenated aliphatics by zero ‐valent iron. Ground water, 1994. 32(6): p. 958-967.
Pankow, J.F. and J.A. Cherry, Dense chlorinatedsolvents and other DNAPLs in groundwater:
History, behavior, and remediation. 1996.

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Tratnyek, P.G. and R.L. Johnson, Nanotechnologies
for environmental cleanup. Nano today, 2006. 1(2):p. 44-48.
Hodson, M.E., The need for sustainable soil
remediation. Elements, 2010. 6(6): p. 363-368.
Nowack, B. and T.D. Bucheli, Occurrence, behavior
and effects of nanoparticles in the environment.
Environmental Pollution, 2007. 150(1): p. 5-22.
Zhang, W., Nanoscale iron particles forenvironmental remediation: An overview. Journal of
nanoparticle Research, 2003. 5(3): p. 323-332.
Mueller, N.C. and B. Nowack, Nanoparticles for
remediation: Solving big problems with little
particles. Elements, 2010. 6(6): p. 395-400.
Chang, M.C., et al., Using nanoscale zero-valentiron for the remediation of polycyclic aromatichydrocarbons contaminated soil. Journal of the Air& Waste Management Association, 2005. 55(8): p.
1200-1207.
CHANG, M.C., et al., Remediation of soil
contaminated with pyrene using ground nanoscalezero-valent iron. Journal of the Air & Waste
Management Association, 2007. 57(2): p. 221-227.
Varanasi, P., A. Fullana, and S. Sidhu, Remediationof PCB contaminated soils using ironnano-particles. Chemosphere, 2007. 66(6): p.
1031-1038.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

A Multi-parameter Water Quality Analysis by Onsite Filtered 2mL
Sample to Monitor Urban River Eutrophication
Akira Kikuchi1,2, Nor Eman Ismail2, Narges Janalizadeh1,2, Musa Mutah1,2, and Muhamad Faiz2,3Institute of Environmental and Water Resource Management, 2Faculty of Civil Engineering, Universiti Teknologi
Malaysia, 3SAJ Holdings SDN. BHD. Johor Malaysia

1

ABSTRACTA new approach for ecosystem level water quality prospecting was examined. Full-automated Ion Chromatographysystem (TOSOH IC2010) was modified for integration with Flow Injection Analysis. 13 of 2mL samples were collected
after onsite filtration by 0.45um syringe filter from river in UTM Johor campus, representing common human impacted
river environment, on 24 February 2012. Then four analytical modes were applied for standard solutions and theenvironmental samples. The detection limit that was signal-noise ratio three in µM/L were as follows, SO 4 2- (0.31),
Cl- (0.50), NO 3 - (0.89), Na+ (0.52), NH 4 + (0.94), K+ (1.24), Mg2+ (0.76), Ca2+ (1.75), NO 2 - (0.02), and PO 4 3- (0.12), andits EC was 0.03 µS/cm. Comparing these values with water qualities of environmental samples, our analytical approach
was satisfactory. The onsite filtration drastically improved the sampling process as simple, and µM/L level analyses
were obtained by only one equipment with 2mL sample. This analysis can be extend to COD, total nitrogen/phosphorous, and etc, which will contribute strategic environmental assessment due to rapid delivered, low-cost, andintegrated data.
1. INTRODUCTION
Environmental assessment has multiple scheme [1][2]. The
assessments for environmental acts implementation in thehigher planning processes than project level is particularly
categorized as strategic environmental assessment (SEA) [1].
Then, the conventional environmental impact assessment(EIA) functions under the SEA at project level [2]. It is clear
from the viewpoint of environmental decision- making
processes, in the sense that as one move down the hierarchy
from policies, plans, program (PPPs), and then to project, thenature of decision-making has prescribed specifications [2].
There in the water matters, standard methods for water
quality analysis [3] have been main technical procedure for
EIA under such object specified conditions. On the other
hand, even though SEA at higher PPPs process has uniquefunction to enhance more useful problem prospecting andeffective environmental programs-projects development[1][2], however the methodologies for suitable water quality
monitoring for SEA are still on a development process [4][5].
Hence, technical development for strategic water quality
assessment for SEA has been focused on in this study.
Elevated levels of phosphorous in conjunction withnitrogen can negatively affect aquatic ecosystems in many
ways in terrestrial fresh water bodies [6][7]. The increased
growth of algae and aquatic macrophytes and distinct shifts inspecies composition are a prominent symptom ofeutrophication [8]. These blooms generally contribute to awide range of water quality problems including summer fishkills, foul odours, and tainted drinking water. Furthermore,certain cyanobacteria produce and release toxins that can kill
livestock and may pose serious health threat to humans. To
monitor phosphorous and nitrogen concentration ofenvironmental water is important [6][7]. From another view
point in general environmental water quality monitoring, themeasurement of common anions (SO 4 2-, Cl-, and NO 3 -) andcations (Na+, K+, NH4+, Mg2+, and Ca2+) found inenvironmental water is essential for some environmental
waters including rain, lake, underground, and river water [9].
Ion Chromatography (IC) have been recognized to be very
useful approach for the determination of anions and cations,and Flow Injection Analysis (FIA) is also very fundamental
for environmental chemistry. Herein an integration of IC and
FIA was examined on analysis for phosphorus, inorganic
nitrogen ions, other common anions and cations analysis by
only one vial of 2mL onsite filtered sample as a convenient
multi-parameter water quality analysis.
2. Experimental
2.1 IC-FIA integration
IC-FIA integration was performed on a TOSOH model
IC2010 equipped with the following devices: vacuum
degasser, double plunger pump that capacity is 6uL x 2,
column oven, auto sampler of capacity 150-300 samples, andconductivity detector (TOSOH, Tokyo, Japan). The
integration with FIA was proceeded by modification of
merging process of double plunger pump by installing a 1-2, 3
switching valve (Fig. 1). Applying 1-2 position of theswitching valve, eluent from double plunger pumps weremerged before sample injector as conventional IC system.
Then, applying 1-3 position of the switching valve, a plunger
pump sent carrier and a sample segment was injected in the
carrier flowing, and reagent sent by the other plunger pump
was merged after the sample injector. Whereas, injected
sample and reagent were merged for every 6uL alternately.

786

0.1Mol/LHClasreagent
solutionA,and
N-(1-Naphthyl)ethylene diamine dihydrochloride (0.05g)
was dissolved to 50mL ultra pure water as reagent solution B.
The reagent solution A and B are mixed just before analysis
as coloring reagent for FIA to analyze NO 2 - [11]. Flow rate
was 0.25 mL/min x 2, injection volume was 50uL, andduration of an analysis was 3min.
PO 4 3- analysis: Stock solution of KH 2 PO 4 (0.06804g) was
solved to 50ml ultra pure water as 0.01 M/L stock solution,and mixed with the ultra pure water for calibration purpose
for PO 4 3- analysis. Ammonium molybdate (0.274g),
antimony potassium tartrate (0.0125g), concentrated sulfuric
acid (3.35mL), L-ascorbic acid (1.5g), and sodium
dodecylsulfate (0.05g) were dissolved to 50mL ultra purewater as coloring reagent [12]. Flow rate was 0.5 mL/min x 2,
injection volume was 50uL, and duration of an analysis was
3min.
EC analysis: 0.1 mM/L KCl was used for 147uS/cm
standard solution for EC at 40oC. Flow rate was 1.0 mL/min,injection volume was 50uL, and duration of an analysis was
3min.

Fig. 1. Schematic depiction of IC and FIA system, and its
integration by a switching valve.

2.2 Manifold for IC and FIA analysis
Using, IC mode that was by 1-3 position of switching valve,
simultaneous anions and cations separation was carried out
using TSK gel Super-IC-A/C (3 um particle size, 0.2 mequiv./
ml cationic exchange capacity, 150 x 6.0mm I.D.) with
conductivity detector. Using FIA mode that was by 1-2
position of switching valve, two channels FIA manifolds
were carried out, in which, 2m SAS tube was applied forreaction at 40 oC for EC analysis, and 5.5m and 4.0m PTFE
tube were applied for reaction at 40 oC for NO 3 - and PO 4 3analysis, with 540nm and 660nm spectrophorometric detector,
respectively.
2.3 Chemicals and analytical conditions
All reagents used in this study were analytical-reagent gradepurchased from Sigma Aldrich and Merck, and ultra purewater system (Millipore Direct Q3) was used to prepare
standard solutions, washing solvents and eluents. Analyticalbalance, ATY 224 (Shimadzu, Tokyo, Japan) was used tobalance reagents.
Common Anion and Cation analysis: A stock solution of
Na 2 SO 4 , KNO 3 , NH 4 NO 3 , and MgSO 4 , and a different
CaCl 2 solution were solved to 0.1M/L, respectively. These
two stock solutions were mixed with the ultra pure water forcalibration purpose for common anions and cations
simultaneous analysis. 10m M/L tartaric acid with 3m M/L
18-C-6 was prepared for eluents [10]. Flow rate was 1.0
mL/min, injection volume was 50uL, and duration of ananalysis was 7.5min.
NO 2 - analysis: Stock solution of NaNO 2 (0.069g) was
solved to 50mL ultra pure water as 20mM/L stock solution,and mixed with the ultra pure water for calibration purpose
for NO 2 - analysis. Concentrated hydrochloric acid (12M/L,
417uL) was dissolved to 50mL ultra pure water as 0.1M/L
HCl solution. Sulfanil amide (0.5g) was dissolved to 50mL

2.4 Field sampling (add onsite filtration)
13 environmental water samples were collected into each of
2mL plastic vials by onsite filtration (0.45 um nylon filter).
Sampling locations were where groundwater discharging
from a cut slope beside a road (S1), forested upstream (S2-S3)
and its downstream of developed campus area (S4-S6), anddischarge from the campus (S7), and oxidation pond (S8-S13)
that discharge to river between S5 and S6 in UTM Johor
campus. These are representing human impacted water
environment. The sampling was performed of 12:00 to 13:00
on 24 Feb. 2012.
3 RESULTS
Calibration graphs were obtained by plotting peak area
against the concentration of standard solutions. Linear
calibration graphs (r2>0.999) were obtained for anions andcations simultaneous analysis in the concentration range
0-0.8mM for common anions (SO 4 2-, Cl-, and NO 3 -),0-0.4mM for cations (Na+, NH 4 +, K+, Mg2+, and Ca2+),0.0002-0.025mM for NO 2 -, 0.00005-0.05mM for PO 4 3-, and
0.005-2.0mM (7.35-2940 uS/cm) for KCl.
The detection limits determined at a signal-to-noise ratio of
three were 0.31 uM/L for SO 4 2- (29.5 ppb), 0.50 uM/L for
Cl- (17.7 ppb), 0.89 uM/L for NO 3 - (12.5 ppb), 0.52 uM/L for
Na+ (11.9 ppb), 0.94 uM/L for NH 4 + (13.2 ppb), 1.24 uM/L for
K+ (48.5 ppb), 0.76 uM/L for Mg2+ (18.4 ppb), 1.75 uM/L for
Ca2+ (70.0 ppb), 0.02 uM/L for NO 2 - (0.7 ppb), 0.12 uM/L for
PO 4 3- (11.4 ppb), and 0.03 uS/cm for EC (Table 1).
As shown in Table 1, the water quality of stream water
were ranging 4.6-213.4 uM/L for SO 4 2-, 68.6-689 uM/L for
Cl-, 3.0-19.0 uM/L for NO 3 -, 79.1-592.8 uM/L for Na+,
1.2-416.5 uM/L for NH 4 +, 3.8-91.9 uM/L for K+, 32.9-96.2
uM/L for Mg2+, 61.6-360.4 uM/L for Ca2+, 0.2-2.4 uM/L for
NO 2 - , 0.2-22.4 uM/L for PO 4 3-, and 292.4-2490.9 uS/cm for787

Table 1. Water quality of stream water in Univeristy Teknologi M alaysia on 24 Feb. 2012.
S1
S2
S3
S4
S5
S6
S7
S8
S9
S10
S11
S12
S13

Groundwater discharging from cut slope
Natural stream in forested area
Nagural stream after small human effect
Up stream, open waterRiver, upstram of oxdation pond
Downstream of oxdation pond
River, infront of security gate
Oxdation pond (NE)
Discharge from oxdation pond
Oxdation pond (NW)
Oxdation pond (SE)Oxdation pond (SW)
Inflow water to oxdation pond

SO42[uM/L ]
14.7
4.6
15.3
14.1
36.9
108.5
110.2
160.4
157.2
155.3
167.0
174.6
213.4

Cl
[uM/L ]
124.1
68.6
93.5
77.4
109.5
287.9
311.0
404.3
428.8
454.1
481.9
517.1
689.0

NO3[uM/L ]
18.8
6.3
15.4
8.8
3.0
9.9
19.0
9.2
11.5
6.4
6.8
7.9
12.9

Na
[uM/L ]
111.9
79.1
105.8
93.7
130.2
259.0
263.3
318.0
352.5
371.3
379.7
404.6
592.8

NH4+
[uM/L ]
3.6
2.6
2.7
2.6
1.2
119.3
111.8
143.4
235.9
227.0
255.6
360.5
416.5

K+
[uM/L ]
3.8
5.7
8.5
7.5
17.8
43.5
50.4
58.5
61.4
65.7
70.1
78.4
91.9

Mg2+
[uM/L ]
45.3
52.5
45.4
96.2
33.6
36.3
37.3
40.0
37.6
34.9
32.9
37.8
40.6

Ca2+
[uM/L ]
61.6
72.5
79.4
165.6
200.7
292.1
294.4
315.9
327.9
314.1
315.5
352.9
360.4

NO2[uM/L ]
2.4
0.2
0.8
0.2
0.2
0.9
1.6
2.0
0.9
0.5
0.2
0.2
1.7

PO43[uM/L ]
0.190
0.174
0.183
0.207
0.168
0.280
0.620
0.348
0.504
0.541
0.696
7.340
22.364EC[uS/cm ]
344.9
292.4
358.2
554.1
594.6
1166.61206.01451.81641.21610.51720.72022.12490.9Environmental sample data (max )Environmental sample data (min )

213.4
4.6

689.0
68.6

19.0
3.0

592.8
79.1

416.5
1.2

91.9
3.8

96.2
32.9

360.4
61.6

2.4
0.2

22.4
0.2

2490.9292.4

Caribration graph range (max )Caribration graph range (min )

400
0

400
0

400
0

400
0

400
0

400
0

400
0

400
0

25
0.2

50
0.05

2940
7.4

Detection Limit*

0.31

0.5

0.89

0.52

0.94

1.24

0.76

1.75

0.02

0.12

0.03

*

-

: The detection limits determined at a signal-to-noise ratio of three.

[uM/L]
250
200
150
100
50
0

SO42-

400
300

M g2+

[uM/L]
800-[uM/L]
100
NO380
60
40
20
0
0
20
-

Cl

600
400
200
0
400

Ca2+

300

15

200

200

10

100

100

5

0

0

0

0

+

5

10

Site number

15

0

5

10

Site number

15[uM/L][uM/L]

700
500

400
300
200
100
0

300
100
0
25
20
15
10
5
0

NO2

0

Na+

5

10

15

Site number

PO43-[uM/L]
100
K+
80
60
40
20

NH4+

PO43-

1.60

3000

1.20 (large scale)
0.80
0.40

1000

0.00
0

5

10

Site number

15

0

5

EC[uS/cm]

2000

10

Site number

15

0
0

5

10

Site number

15

Fig. 2 Water quality of stream water in Universiti Teknologi malaysia on 24 Feb. 2012. Comparing todetection limit shown in table 1, the resolution among analyte anions, cations, and EC were quite satisfactory,
and the methods were useful for the accurate determination of these parameters in common river water.

EC (Table 1). Comparing to detection limit shown in Table 1,
the resolution among analyte anions, cations, and EC weresatisfactory in order to assess the trend of environmental
water quality difference in UTM campus (Fig. 2).
4 DISCUSSION
In this study, common anions and cations simultaneous
analysis for SO 4 2-, Cl-, NO 3 -, Na+, NH 4 +, K+, Mg2+, andCa2+ were performed by IC mode. Then the analyticalresolution was satisfactory and can be considered useful forthe accurate determination of these common anions andcations. Moreover, the analysis is a part of an advanced
non-suppressor type IC by “polyacrylate– based weakly
acidic cation-exchange resin column” with weak-acid eluent[13][14]. By the analytical column, a series of very functional
water quality analysis are obtained, such as, these 8 ions plus
PO 4 3and
SiO 4 2simultaneous
analysis[15],HCO 3 analysis[16], simultaneous heavy metal analysis for
Ni2+, Zn2+, Co2+, Mn2+, and Cd2+ [17].
Applying thisadvanced IC, there is a potential to develop integrated IC forcommon cation and anion, nutrients, and heavy metals by
only one analytical column and equipment.
However, some important parameters, for example COD,
total nitrogen (TN), total phosphorous (TP) are impossible toanalyze by conventional IC, and generally seawater samples
are not applicable for IC. In addition, some parameters, forexample IC for phosphorous has low sensitivity forenvironmental monitoring [15]. On the contrary, FIA has

advantage for these analyses [11][12][18][19]. Accordingly,
because of similarity for system design between IC and FIA,
these analytical systems were integrated in this study (Fig. 1).
Consequently, SO 4 2-, Cl-, NO 3 -, Na+, NH 4 +, K+, Mg2+, Ca2+,
NO 2 -, PO 4 3-, and EC were able to be analyzed from only 2mL
onsite filtered sample in a plastic vial. This multi-parameter
analysis can be extended to TP [12], TN [18], COD [19],HCO 3 - [16] within another one 2mL vial sample, easily, andthe analytical capacity is several hundred samples regarding
the capacity of auto sampler of IC2010.
In order to embark on effective environmental restoration
programs and projects, SEA is quite important [1], where new
methodologies for SEA need to be developed [8][9].
According to literature [8][9], expected properties of suchnew approach is “rapid delivery of results, on-site, low-cost,and capacity to acquire large number of observations within ashort time frame”. Regarding the requirement, the concept ofon-site analysis was eliminated, and time frame for rapid
delivery of result is assumed as one day, and then, convenientexperiment procedure, satisfactory accuracy, capacity of
automated large number sample analysis, and integrated
multi-parameter analysis were of more concern in this study.
According to our analytical approach, because of reduction ofsample volume, even concept of onsite analysis waseliminated, field sampling process and sample transportation
process were drastically improved, and then, convenient andenough accuracy of multi-parameter large number sample
was available, which will contribute to SEA as a new788

technical approach that improves problem recognition
processes in environmental prospective and profiling.
In water environmental management, the communicationgap (CG) between water authorities and public causes
difficult chronic management problem in water-environment
monitoring and assessment [5]. The potential solution is toprepare useful information, where our multi-parameter
convenient approach will contribute to give new approach tosolve the CG by rapid delivery of data, low-cost, and highcapacity to acquire large number of observations of integrated
data.
5. CONCLUSION
Ion Chromatography and Flow Injection Analysis integration
was performed, which was satisfactory to analyze SO 4 2-, Cl-,NO 3 -, Na+, NH 4 +, K+, Mg2+, Ca2+, NO 2 -, and PO 4 3- in uM/Llevel, and EC in s 0.03 µS/cm by only one equipment within
2mL onsite filtered sample. The analysis can be extended to
COD, TP, and TN. A convenient water quality prospecting
was available for strategic environmental monitoring due torapid delivered, low-cost, and integrated data.
6. ACKNOWLEDGMENT
This study was supported by FRGS (4F089) and ERGS
(4L015) fund by Ministry of Higher Education, Malaysia. We
express our gratitude to TOSOH for all the support for thisstudy.
7. REFERENCES[1] Gauthier M, Simard L, Waaub JP, “Public participationin strategic environmental assessment (SEA): Criticalreview and the Quebec (Canada) approach,” Env. Impact
Assessment Review, 2011, vol.31, pp48-60.
[2] Fischer BT, Wood CJC, “Policy, plan, and program
environmental assessment in England, the Netherlands,
and Germany: practice and prospects,” Env. andplanning, 2002, vol.29, pp.159-172.
[3] Clesceri LS, Greenberg AE, Eaton AD, “Standard
methods for the examination of water and wastewater”
20th ed. American Public Health Association,
Washington, 1998, 1325 p.
[4] Graveline N, Maton L, Luckge H, Rouillard J, Strosser P,
Palkaniete K, Rinaudo JD, Taverne D Interviews E,“Potential uses and constraints of emerging water quality
Monitoring tools: an operational perspective,” Trends inAnalytical Chemistry, 2010, vol.29, pp.378-384.
[5] Kikuchi A, Hakim L, Heryansyah A, Romaidi R“Significance of the Easy-to-use Water Quality Checker
for Participative Environmental Monitoring andExperience Based Learning,” Journal of Tropical Life
Science, 2010, vol.1, pp17-21.
[6] Smith VH, Tilman GD, Nekola JC, “Eutrophication:
impacts of excess nutrient inputs on freshwater, marine,
and terrestrial ecosystems,” Env. Pollution, vol. 100,
1999, pp.179-196.
[7] Elser JJ, Bracken MES, Cleland EE, Gruner DS, Harpole
WS, Hillebrand H, Ngai JT, Seabloom RW, Shurin JB,
and Smith JE, “Global analysis of nitrogen andphosphorus limitation of primary producers infreshwater, marine and terrestrial ecosystems,” Ecology
Letters, 2007, vol. 10, pp.1135–1142.
[8] Carpenter SR, Caraco NF, Correll DL, Howarth RW,
Sharpley AN and Smith VH, “Nonpoint pollution of
surface waters with phosphorus and nitrogen,”
Ecological Applications, 1998, vol. 8, pp.559-568.
[9] Stumm W, Mrgan JJ, “Aquatic Chemistry; an
Introduction Emphasizing Chemical Equilibria in
Natural Water,” 2nd ed., Wiley, New York, 1981, pp.780.
[10] Masuda W, Tanaka K, Goto R, Hasebe K, Kikuchi A andNakagoshi
N,“Ion-exclusion/cation-exchange
chromatography for water quality monitoring of riverwaters,” J. Japan Industrial Water Association, 2007, vol.
585, pp.80-86.
[11] Higuchi K, Motomizu S, “Flow-Injection Sepectrophotometric Determination of Nitrite and Nitrate in
Biological Samples,” Analytical Science, 1999, vol. 15,
pp129-134.
[12] Higuchi K, Tamanouchi H, Motomizu S, “On-line
Photo-Oxidative Decomposition of Phosphorous
Compounds to Orthophosphate and Its Application to
Flow Injection Spectrophotometric Determinations of
Total Phosphorus in River and Waste Waters,”
Analytical Sciences, vol. 14, 1998, pp941-946.
[13] Tanaka K, Ohta K, Fritz JS, Matsushita S Miyanaga A,“Simultaneous ion-exchange chromatography with
conductimetric detection of anions and cations in acidrain waters,” J. Chromatography A, 1994, vol. 671,
pp.239-248.
[14] Mori M, Tanaka K, Helaleh MIH, Xu Q, Ikedo M, Ogura
Y, Sato S, Hu W, Haebe K Haddad PR, “High-speed
simultaneous ion-exchange chromatography of anions
and cations on a weakly acidic cation-exchange resin
column,” J. Chromatography A, 2003, vol. 997,
pp219-224.
[15] Nakatani N, Kozaki D, Masuda W, Nakagoshi N, Hasebe
K, Mori M, Tanaka K “Simultaneous spectrophotometricdetermination of phosphate and silicate ions in riverwaterby usin ion-exchange chromatographyic separationand post-column derivatization,” Analytical Chimica
Acta, 2008, vol. 619, pp110-114.
[16] Mori K, Ikedo M, Hu W, Helaleh MIH, Xu Q, Itabashi H
Tanaka K, “High-speed ion-exchange chromatographyof dissolved carbon dioxide on a small weakly acidic
cation-exchange resin column with ion-exchangeenhancement columns of conductivity detection,” J. of
Chromatography A, 2005, vol. 1092, pp250-253.
[17] Mori K Tanaka K, “Separation of Heavy Metal
Ion(Ch.2.5),” The practice of ion chromatography,
Tanaka K, Ed. Mimizukusya, Tokyo, Japan. 2009, p228.
[18] Kitami H, Ishihara Y, “Simultaneous Dtermination of
Total Nitrogen and Total Phosphorous in Sewage Water
by Flow Injection Analysis after Continuous Flow
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789

[19] Vallejo-Pecharroman B, Izquierdo-Reina A, de Castro
MDL, “Flow injection determination of chemical oxgen
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pp.1261-1264.

790

THE INTELLIGENT PERFORMANCE INSTALLATIONS IN ENERGYEFFICIENCY IN GREEN BUILDINGS
Hooman Abadi, Rouzbeh Abadi, Poona Abadi
Universiti Kuala Lumpur (Unikl) /Business school, Kuala Lumpur, Malaysia,
Universiti Teknologi Malaysia (UTM), Faculty of Civil Engineering, Kuala Lumpur, Malaysia,
Universiti Teknologi Malaysia (UTM), Faculty of Management, Kuala Lumpur, Malaysia
hooman_abadi@yahoo.com
rouzbeh.abadi@me.com
abadi.poona@gmail.com
ABSTRACT:
Increasing population, growth and diversity of the people demands in the work and communal activity and living
issues like achieving to the participated aims, move welfare, change the life culture and public in structure and or
expansion of the building's foundation level and or possibility to enjoy of the day and fresh technology in the life
brought out the unused issues in the different fields such as safe, protection, repair and maintain of the instruments
and installations, presenting the favorable service, suitable consumption and consumption optimization in thevarious energies
in collection. Remedies about the protection issues in scattered and some had acceptable solutions, Supervisionon good performance and leading of installations and instruments in order to the suitable exploitation and the movesuitable service presentation has been applied the traditional methods and the limited solutions with significant cost.
But it needed do the abundant effective practical measures to apply the modern management in order to the suitable
exploitation of a building in generally such as useful age of installations and instruments, in time service andmaintenance, optimum consumption of kinds of energy and determining at least cost in any country.
It can be said in courage that is needed a basic turn to the views. Perhaps, there are fewer green buildings in theworld that presents a statistic of the energy consumptions, their wastes, to determine the costs based on statistics ofthem.
In the present paper management techniques and integrated quick control equipment while define
the mechanical and electrical installations and instruments coordinated clever management system and this system.
How has efficiency a powerful instrument to manage a building (construction)
Keywords: Intelligent, Performance Installations, Energy Efficiency, Green Buildings
INTRODUCTION
Increasing energy consumption and as a result
increasing prices during the current decade caused manyenergy consumers, and even retail consumers try
to reduce consumption, and it results from various kindsof savings. Meanwhile, some scholars believe that
increasing price of energy as a driving factor forsavings, although bio environmental damages resulted
from excessive consumption of energy caused toconsume wisely too. For answering new requirements
such as lower consumption and more efficiency, noticeand resorting to innovation and inventions and utilizing
new technologies in all sections of production, as wellas transportation and distribution of energy, and also in
consumption section is unavoidable. Therefore, formanaging and steering a building and dealing with
events with the lowest unsuitable effect for presenting

proper services need of changing view and exploiting
management methods from traditional and manual
methods to contemporary ones can be felt clearly.
Because humane based on their physiological structure
and limited capacities in time of unacceptable events arevulnerable and will lose their sagacity and mental
control for resisting or if controlling or supervision
domain is vast, negative effects of tediousness or lackof concentration will occur. accordingly, new modern
machineries as a type of movement with the goalof subsiding humane force with mechanical andelectronic ones should be utilized. Geographicalsituation of Iran from one hand and necessity of savingin costs and avoiding wasteful and maintaining national
capital such as water resources, energy, electricity,
repairs, precautionary service of equipment andimportant huge building infrastructures such as health792

centers from the other hand demands this changeof method in country level, which has stated below:
1- Relying on sectional scientific specialists andtopical technical knowledge
2- Designing and building required mechanicalequipment inside the country (as much as
possible)
3- Having services and customer services after
implementation
This movement for building exploitation managementfrom the manual system to an effective electronicsystem with an ability of integration in further stages isfrom inside the country and with relying on all of local
specialist forces and immediate facilities than with aproper investment and along with a codified and assuredplan, this movement for buildings of the country will beachieved its goals. It is believed that considering current
capabilities and facilities and with utilizing them andinfrastructures with utilizing the intelligent management
system, this national capital will be maintained and savein cost and also functionality of equipment will becoordinated and competent.
A requirement of Management of a building:
Each building and especially huge buildings such as tallor wide in ground level regarding their functional
requirements and number of infrastructures andequipment presented in them have some requirements inmanaging and proper and efficient exploitation of themwhich some of them are as follows:• Reducing power consumption without creatinga lack of services• Lack of presence of identity and repair recordof infrastructures and equipment and numberof services• Lack of precautionary and on time services forpresent equipment• Any kind of setting or control is done manuallyby operators, and no automatic setting
equipment is using. Therefore, settings are
fully depended on technical knowledge andpersonal preferences of the operator not toa systematic knowledge and what should bedone technically.
• High cost of human-resource
wage for
operating and maintaining equipment• High rate of depreciation because lackof the intelligent system as a tool for
maintainer's forces and operators forcontrolling functional quality• Lack of presence of a precautionary
maintenance program (P.M) for equipment
and building infrastructures• Planning with saving goals in cost of electricity
power, fossil fuel, etc. is not functional

because of lack of required information andneeded equipment that its result is various
high costs and undesirable and inefficientresults.
• Really accurate value of activity of backup
services manager is not clear because of lackof information about the way they work andnot being familiar with their facilities andmethod of their reactions to needs.
• Lack of using expert specialists for presenting
proper technical services to buildings,
especially in deprived regions
To the above issues, controlling another equipment
including always on and non-always on such as startingand stopping elevators according to requirements,
opening and closing special doors such as emergency
doors in time of disasters and such these will be added.
Necessity of answering these needs for desirable
managing a building caused to invent electronicsystematic solutions that are defined for a specific job
which efficient management of buildings will bepossible with relying on these tools.
What is the meaning of intelligent management ofa building?
One of the tasks of management is evaluating deviancefrom the goal and guiding the set in each time periodintervals so the set under management is moving toward
goals and achieve its goal. Building management withthe aim of presenting on time and desirable services andwith optimum cost is not separated from this subject
too. For this reason, some tools are needed including
gathering information from equipment and installationsof building in form of signals such as temperatureof warm water, the amount of consumed cold water,
being in ON or OFF state of equipment, pumps andsuch as these installations. Being in move or stopped
stateof facilities such as elevators or some
informationof the security system, lighting
system, Coccus and, etc. and then this gathered
information is placed for analysis process that can behandled by one or more persons manually or throughelectronic systems and by computer automatically. It
seems that regarding
the number of signals
and number of different requirements of exploitationand capabilities
of electronic machines
and present today microprocessor equipped machines
such as computers, analyzing these signals
automatically is unavoidable.
In this situation,
all the varying possibility is predicted by functional
software or a desirable algorithm, logical continuity ofcases observed and in case of mistakes,
awarning will be issued or another decision will be madean issue order to implement that.

793

Results of analysis prepared in form of made decisions
ready for implementation, and managerial order will
issue. These orders and commands can be in various
digital forms such as make consumer keys on or off
such as lighting keys, starting pumps and electrometeror includes combined and group functions such as
stopping all of elevators or opening all emergency
doors. Implementing decisions too can be granted toa person or a group, or it can be granted to somemachines because most of the information is gathered
during 24-hour day and night, therefore, managementof a building is like management of a factory that is
running 24/7. In addition to managing daily routines,
another management requirement is observing
functional in building in time, day, month, season andyear, which are relying on the same gathered
information that preparing statistics, generating
periodical reports, and codifying proper algorithm for
operation of equipment is a part of this task. Such thissystem, including, transmitting and receiving signals,
commands, reports to remote location and having
records of operations in an active archive called
intelligent management of building, which is consistedof hardware, software, and controlling systems.
Capabilities and technical advantages of Intelligent
Management of building:Once such these powerful and efficient tools areprovided for managing a building, in addition tohabitual and communal jobs in building management
modern capabilities can be achieved, which were
impossible, until now, while even for doing customary
and shared tasks too, unused definitions and capabilities
will be stated that below points can be mentioned:
Collect, register and record all input data:
Input data are analog and digital signals that transmit all
required information about situation and stateof equipment and installations such as determining
energy's consumption rate, determining the excessiveor shortcoming amount of consumption comparedto factors, temperature of water, flow of water, leakage
in pipes, voltage and amperage and power consumptionof a machine or a set of consumer machines, being inon or off state of an electrometer, state of an elevator ineach moment of time, or a door or window being open
or close, and such these data sent to the system. With
collecting data, record and analyzing and processing
them, diagrams and process of their changes in a desired
period of time can be prepared, viewed or study
or transmitted for a specific reason to a remote location.
Controlling decision making based on collected data:
Recorded data of quantities compared to a source
quantity, for example, temperature of received warm
water compared to desired temperature, and decision
will be made based on it and if water is warm enoughheater will be turned off. Generally such thesecontrolling and commands are done on installations but
functional history of parameters and mechanismof control will not be recorded. Moreover, any decision
about controlling circuit should be done in the location
which equipment is installed while in the intelligent
management system of building change order inreference quantity's values (set points) or mechanismof control (such as PID or PI) or operation command
can be issued from command room and utilize software
flexibility in the system too.
It coordinated and on time operation in the building:In buildings function of equipment and installations
traditionally grouped into various groups, for example,
control and operating powerhouse is grated to a groupand repairing and supervising power systems is granted
to other groups and maintain and securing is granted tosecurities, but integrity and coordination between
functions such as coordination of job of a fire alarmsystem with air-conditioning system and elevators forbuilding management is very desirable and haveimportance. This coordination will lead to a majorsaving in costs, energy consumption, and depreciation.
Expanding managerial services in form of the
management information system:
Increasing buildings in country and need of technicalspecialist for control, operate, service, and maintaining
equipment and installations from one hand and lack of
the possibility of utilizing required specialists,
especially in remote areas from other hand statenecessities
of intelligent management ofbuilding so with this method it will be possible totransmit registered and recorded information throughtelecommunication networks to users and experts inremote location far from building hence management ofbuilding both technical and non-technical relying ontransmitted data in form of management information
system (MIS) and based on real statistics can makeimportant decisions that lead to major savings in
consumption of water, electricity, fuel, depreciation, andavoid lack of coordination.
Scheduling and planning for service and repair:
One of the weaknesses that there is in maintaining,
servicing and operating of technical systems of buildingin the country, including industrial and non-industrial isthat preparing repair history and scheduling for repair
and precautionary maintenance (PM), if not consideredas impossible but is very difficult and cost lots of money
and is inactive. With the intelligent management
system, it is possible to record functional history ofequipment and utilize it in the future. This matter,
especially when it comes to ordering for spare parts,
warehouse handling, and reducing costs is very
important. With the help of the intelligent management
system, it is possible to schedule for periodical repairs
and determine current or yearly repairs, and with the794possibility of necessary predictions reduce most of thecosts and also for precautionary service andmaintenance by the system it will be possible to fightagainst majority of dangers and prevent loss.
Increasing useful life span of equipment:
Implementing ordered plan and precautionary service
and maintenance will cause to determine working hoursof each machine in advance, identity card of eachmachine, including record of all past problems, andspare parts replaced should be prepared and througha software program service and maintain. Implementing
these steps will lead to decreasing depreciation andenergy consumption and will increase the useful life
span of equipment and make them always stand by andready to work.
Savings:
Getting statistics of consumption of water, electricity,
gas or diesel is the first step for finding a solution forsaving that with the help of intelligent management suchas these statistics is provided easily. And based on it ispossible to calculate what type of saving in which partcan be done. These calculations are simply done bycomputers and cost of any type of control for each partis determined. Managing power distribution system,
making lighting system intelligent, controlling airconditioning system with observance of saving energy,
and efficient managing of warm water installations andradiators or chillers after implementing parametersof savings are all possible with the help of the intelligent
management system.
For example, it is reported that in a tall building in
United States between 20 to 41% in energy cost for each
unit of infrastructure saved. Moreover, according tovarious reports, including presented information in table
no.1, recorded statistics showed definitive savings after
utilization of the intelligent management system in all
variouseconomic
sections,
includingurban
installations.
Economic effects of the intelligent management system:
Because of various utilizations of buildings, servicing,
maintaining, and keep all equipment and installations
including power, mechanical, and low pressure inbuildings ready to work has special importance andnecessity of on time and proper service to residents andguests is clear. With a glance to maintaining cost ofbuilding and installations and also fuel, electricity,
lighting of some buildings including health centers, itwill be determined that while presenting proper services
for stopping increasing mentioned case's costs or evendecreasing them a solution should be thought.
Equipping buildings with the intelligent management
system along with technical and efficient effects of thatwhich has mentioned because of creating the properbasis for creating job opportunity in technical level andcurrent technology for university graduates andtherefore, increasing technical level of services andmaintaining of buildings from one hand and service,
maintain, and consumption of fuel, electricity andenergy from the other hand bring suitable economic
effects. Moreover, implementation level of this systemin building with a scheduled plan in a few years can be
upgraded and utilization of system and level of savings
will be increased.

CONCLUSION:
Presenting suitable and on time service along witheconomic savings in energy consumption and electricity
energy is one of the first and most important actions inproper management of building. Utilization of buildings
in all over the country raises another case that for
presenting suitable services should find new solutions.
Building management now needs new tools that
intelligent management system is designed and created
for answering these needs. With the help of this system
in addition to common savings, it is possible to manage
different parts of a building under an integrated andcoordinated control and from a remote location
supervise and control buildings and till some extent
diagnose technical difficulties in them. From other hand
this system with presenting managerial services such as
possibility of preparing periodical reports and service
and maintenance reports and such these present special
facilities to the managers of buildings. Valuable
experiences in design, build and implementing thissystem shows that with spending a little cost canachieve noticeable savings.
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Environmental Life Cycle Assessment ofModern Office Buildings, Steel Construction
Institute, Ascot, UK, 1998.
[19] B. Berge, the Ecology of Building Materials,
Architectural Press, Oxford, UK, 2000.
[20] D.J. Clements-Croome, K.G. Jones, G. John,
Through life environmental business modelingfor sustainable architecture, CIBSE
Proceedings of Conference on Building
Sustainability, Value and Profit, Edinburgh,
UK, 2003.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Study the Effect of Polyvinyl Chloride (PVC) Addition on the Permeability ofAnbar Soils
Dr. Ahmed H. Abdul Kareem & Omar Mustafa
Engineering College-Anbar University, Iraq
ABSTRACTThe soil is considered the most abundant building material in the world and after the only local resource available to theengineer for his work. On the other hand, the inclusion of materials such as soil stabilizing agents allowed to improve some ofits properties (mainly strength, permeability), reduce costs and environmental effects, especially if such materials fromindustrial by products. The aim of this research is to study the effect of PVC addition on the permeability of Anbar soils.
Permeability testing for these soils by laboratory tests working ( i.e. the addition of different percentages of PVC, 5%, 10%,15%, and 20%) to four types of fill soils from different areas in Anbar government. Classification tests were done on thesesoils. The addition of PVC to the soils lead to significant decrease in soil permeability and this serves in improving the soil
layers that are placed under buildings and roads as sub-base or sub-grade layers. Also, the study found that the addition of
PVC at least 15% to the soils will reduce the coefficient of permeability, k, by up to 99% and that help to use it instead ofconcrete in the lining of the water treatment plants. The study also found the coefficient of permeability, k decreasesignificantly in soils containing a high proportion of gypsum.
Keywords: Permeability, Soil Stabilize, PVC addition, Anbar soil
especially looking for a good stress-strain behavior andpermeability.
The properties that are attempting to change into soils
through a process of stabilization are: volume stability,
strength, permeability, compressibility and durability.
With respect to soil stabilization methods, the following
have been referenced in literature (April & Castrillon,
1993, Aguirre & Velez, 2005, Amaya et al., 1978,
Caicedo & Herrera, 2004, and Gomez et al., 1983) tostabilize granular sub-bases: mechanical stabilization
throughmixtures,mechanicalstabilizationbycompaction, stabilization and drainage, lime stabilization,
stabilization with lime and fly ash, stabilization withasphalt and cement stabilization [2,3,4,5,6].
The most commonly used to stabilize granular sub baseunder the concrete, asphalt, compaction and lime;
however are currently investigating methods andunconventional materials which stabilize optimal from thestandpoint of technical, economic, and environmental [7].

1. INTRODUCTION
According to Lamb (1999), the soil is considered the mostabundant building material in the world and after the only
local resource available to the engineer for has work in
the case, it is intended that its characteristics meet therequirements demanded by these jobs [1].
The inclusion of materials such as soil stabilizing agents
allowed to improve some of its properties (mainly
strength, permeability), reduce costs and environmental
effects, especially if such materials from industrial-by
products or recycling, and updated the stabilization
techniques, modification or improvement of soils throughthe inclusion of alternative materials.
The concern to stabilize or modify the soil has generated
inventive approaches, although based on conventional
systems make use of additive and processes following thistrend, we present this research which studies the behavior
in terms of coefficient permeability in its natural state,
and the same material added product that until now had
not been used as modifying agent in this such materials
PVC.
The work focuses on the analysis of behavior in terms of
permeability, for being the most commonly used test in
the different soils in earth dam, and reservoir dams.
The purpose of using PVC is improved the coefficient
permeability, k. These additives can be used with avariety of soils help improve their natural soil these
additives depends on the soil treated and the amount ofadditive used.

3. POLYVINYL CHLORIDE (PVC)
Polyvinyl chloride, -(CH 2 -CHCL-) n – is a synthetic
polymer which is obtained by addition polymerization ofvinyl chloride, and is the raw material for preparation of
PVC.
Raw materials which results in the PVC are common salt
and oil or natural gas. Oil or gas ethylene is obtained
through a cracking process, the salt dissolves in water andsubjected to electrolysis to separate the chlorine in it.
Ethylene, 43%, and chlorine by 57%, are then combined
under heat to obtain a gas monomer, vinyl chloride.
Monomer molecules are linked as results of a reaction
know as polymerization. The product of this process is the

2. SOIL STABILIZATIONThe soils are stabilized to modify existing properties
making them able to best comply with their requirements,

797

PVC resin in its virgin, whose appearance is a fine white
powder [8].
Most times the polyvinyl chloride used with additives that
improve their finish, these are: plasticizers (giveflexibility to the polymeric materials), heat stabilizers( toprevent thermal degradation during processing andincrease shelf life), lubricants (help fluidity during
processing and prevent adhesion to metal surfaces), fillers(lower costs) and pigment (give color, opacity andweathering).
Because faculty have mixed resins with various additives,
PVC has been used for various products, ranging from therigid and durable used for drinking water supply andenergy to the flexible and extensible films for foodpackaging.
The construction and infrastructure accounts for 60% of
products made with PVC benefiting the areas of modern
life or to meet human needs
Plastics are also called synthetic resins and are broadlyclassified into two categories; thermosetting resins andthermoplastic resins. The thermosetting resins include
phenol resin and melamine resin, which are thermallyhardened and never become soft again. Thermoplastic
resins include PVC, polyethylene (PE), polystyrene (PS)and polypropylene (PP), which can be re-softened byheating.

Flow of soil water for non-turbulent conditions has been
expressed by Darcy as(1)
ν =k iWhere
i = Hydraulic gradient h/L
k=Coefficient of permeability (or hydraulic conductivity)
as proposed by Darcy, Length/time.
v= Discharge velocity, which is due quantity of waterflowing in unit time through a unit gross section area ofsoil at right angle to the direction of flow.
Table (1) lists typical order of magnitude (exponent of 10)
values for various soils.
Table (1) Order-of-magnitude values for permeability k,
based on description of soil and by the Unified Soil
Classification System, m/s [9].

100
10-2
10-5
10-9 10-11
Clean
Clean
Sand-silt
Claysgravelgravel andmixtures
GW,
sand
SM, SL,GP
mixturesSC
GW, GP
SW, SPGM

4. PVC’s PHYSICAL PROPERTIESPVC, PE, PP and PS are general purpose plastics. The
features of the particular plastic are determined by its
chemical composition and type of molecular
structure(molecular formation: crystalline/amorphous
structure) PVC has an amorphous structure with polar
chlorine atoms in the molecular structure. Having chlorine
atoms and the amorphous molecular structure are
inseparably related. Although plastics seem very similarin the context of daily use, PVC has completely different
features in terms of performance and functions comparedwith olefin plastics which have only carbon and hydrogenatoms in their molecular structures. Chemical stability is acommon feature among substances containing halogens
such as chlorine and fluorine. This applies to PVC resins,
which furthermore possess fire retarding properties,
durability, and oil/chemical resistance.

The quantity of flow q through a cross section of area A is
Volume/time(2)
Q =k i A
Permeability is a measure of the ease in which water canflow through a soil volume. It is one of the mostimportant geotechnical parameters. However, it isprobably the most difficult parameter to determine. In
large part, it controls the strength and deformation
behavior of soils. It directly affects the following:-Quantity of water that will flow toward an excavation-Design of cutoffs beneath dams on permeable
foundations-Design of the clay layer for a landfill liner.
Two tests commonly used in the laboratory to determine k
are the constant-head and falling-head methods. Figure 1
gives the schematic diagrams and the equations used forcomputing k. The falling-head test is usually used for k <10~5 m/s (cohesive soils), and the constant-head test isusually used for cohesionless soils [9].

5. MECHANICAL STABILITY
PVC is a chemically stable material, which shows littlechange in molecular structure, and also exhibits littlechange in its mechanical strength. However, long chain
polymers are viscose astir 1 materials and can be
deformed by continuous application of exterior force,even if the applied force is well below their yield point.

7. ORDINARY METHOD
When the soil k < 10~5 m/s Falling-head permeability test
using the standard compaction-mold permeameter asshown in Figure 3 where both a disassembled device anda test setup using a 100-ml burette. Great care is requiredin assembly to avoid leaks. Use a meter stick to measure
the hydraulic heads, h 1 and h 2, and the coefficient of
permeability could be fined by Appling Eq. (3) with small
hydraulic gradients exercise care that sample does not
drain [10].

6. PERMEABILITY

798

k = 2.303 *

h
aL
log 1
At
h2

(3)

Where :
k= Coefficient of permeability (or hydraulic conductivity)
as proposed by Darcy, Length/time.
a = Inside area of standpipe, cm2
A = Cross - sectional area of specimen, cm2
L = Length of specimen, cm
t = Elapsed time of test, sec
h 1 =Hydraulic head across sample at beginning of test(t=0)
h 2 = Hydraulic head across sample at end of test (t=t test )8. CHARACTERISTIC OF SOILS
Four soils from different areas in Anbar government wereselected for use in the admixture evaluation. The soil
properties were determined in general accordance withASTM and BS. Tests included Grain Size Distribution(ASTM D 422-02), Atterberg Limit testing (ASTM D
4318), compaction test (ASTM D 698), specific gravity(D 854) and gypsum Content, (BS 1377). Material
classifications were made in accordance to the Unified
Soils Classification System (USCS). A summery of thesoil properties determined in the laboratory and thecorresponding standard testing identification is presented
in Fig.(1) and Table (2).

9. TEST PROCEDURE OF PERMEABILITY
Three SM soils, and one soils of type CL-ML were used.
PVC addition was mixed with each of the soils. PVC was
added to the soils at rate of (5, 10, 15, 20) percent by
weight of the dry soil. Each PVC/ soil combination wascompacted to maximum dry unit weight and optimum
moisture content (O.M.C) for each soil. After the prepared
the compacted test samples which used in falling
permeability testing [11].
10. RESULTS AND DISCUSSION
After examination and obtaining the basic results, PVC
has been added to each soil and at different rates and in amanner to bring the weight of the PVC replaced byequivalent weight of the soil then see the changes on soil
properties by comparing the results of permeability before
and after Addition.
10.1 Soil No.1
The addition of PVC into the soil 1 (SM, silty sandgypseous soil) calculated value of k as shown in Fig (2).
From this figure, it can be seen that the addition of PVC
to the soil at different rates lead to a decrease in k values.
The ppercentage decrease in the k value reaches 97%
when the addition of PVC by 20%.
10.2 Soil No.2 and 3
The effect of PVC addition on coefficient of permeability,k, for soil 2 and 3 (silty sand) shown in Fig. (3) and (4).
These figures show that increasing the proportion of
addition of PVC to the soil leads to a decrease of k values
and this tells us in improving the soil layers that are under
the foundations of buildings or roads and prevent the riseof groundwater, which causes them to reduce the shear
strength of soil.The ppercentage decrease in the k value
reach to 99.2% and 99.7% when the PVC addition of 20%

Fig. (1) Results of compaction tests for Table
four Soils.
(2) Material Index Properties
Parameters
Location
Liquid Limit, LL (%)
Plasticity Index,PI (%)
Finer %
Sand %
Gravel %Dry Unit Weight (kN/m3)
Optimum Moisture Content %USCS System
Description of Soil
Specific Gravity (Gs)
Gypsum Content %Soil 1
Soil 2
RamadiWest Ramadi
NL
NL
NP
NP
Particle Size Distribution
27
14.8
73
85.2
0
0
Compaction Test
16.38
17.96
7.2
13.4
SM
SM
799
Silty Sand (Gypseous Soil)
Silty Sand
2.43
2.64
47
8.02

Soil 3
East Ramadi
NL
NP

Soil 4
North Ramadi
18
7

16.6
83.4
0

84.6
15.4
0

18.53
13.4
SM
Silty Sand
2.58
13.75

17.46
16.6
CL-ML
Silty Clay with Sand
2.6
7.74

10.3 Soil No.4
The soil is silty clay with sand (CL-ML). The addition of
PVC will decrease the coefficient of permeability, k, forall addition as shown in Fig. (6). The decrease of thecoefficient of permeability, k, percentage to 99.8% when
the PVC addition of 20%. In Table (3), we can see thefinal results are calculated to effect PVC addition on
coefficient of permeability, k, for different soil.
Nature% 10 PVC% 5 PVC%15 PVC

Nature% 5 PVC

Coefficient of Permeability,k (cm/s) x 10 -5

and 15% for soil2 and 3. Fig. (5) show effect of PVC
addition on coefficient of permeability, k, for soil 1 to 3
respectively.

6.0
5.0
4.0
3.0
2.0
1.0
0.0

% 10 PVC%15 PVC% 20 PVC

5.69

0.11 0.04

0.02 0.08

PVC Addition

Fig. (4) Effect of PVC addition on coefficient of
permeability, k, for soil 3.

% 20 PVC

Coefficient of Permeability,k (cm/s) x 10 -5

3.00
2.020
2.00
1.00

0.203 0.397 0.313

0.00

0.057

PVC addition

Fig. (2) Effect of PVC addition on coefficient of
permeability, k, for soil 1.
% 5 PVC
4.0
2.0
0.0

% 10 PVC%15 PVC% 20 PVC

Fig. (5) Effect of PVC addition on coefficient of
permeability, k, for soil 1 to 3.

3.86

1.640
0.122 0.103

0.0297NaturePVC Additiion

Coefficient of Permeability,k (cm/s) x 10-6

Coefficient of Permeability,k (cm/s) x 10 -5

NatureFig. (3): Effect of PVC addition on coefficient of
permeability, k, for soil2.

% 5 PVC
10.0% 10 PVC%15 PVC% 20 PVC

7.420

5.0
0.0

0.2420.093

0.092 0.016

PVC addition

Fig. (6): Effect of PVC addition on coefficient of
permeability, k, for soil 4.

800

Table (3) Values of permeability coefficient, k, for different soils withdifferent PVC addition and the percentage decreasing.
Type of Soil

Soil 1

Soil 2

Soil 3

Soil 4

PVC addition %Coefficient ofPermeability, k,
cm/s

The Percentage
Decreasing, %

0
5
10
15
20
0
5
10
15
20
0
5
10
15
20
0
5
10
15
20

2.02 x10-5
2.03 x10-6
3.97 x10-6
3.13 x 10-6
5.70 x10-7
3.86 x10-5
1.64 x10-5
1.22 x10-6
1.03 x10-6
2.97 x10-7
5.69 x10-5
1.07 x10-6
4.17 x10-7
2.48 x10-7
7.6 x10-7
7.42 x10-6
2.41 x10-79.28 x10-89.20 x10-8
1.66 x10-8

0.00
90.0
80.0
84.5
97.0
0.00
57.5
96.8
97.3
99.2
0.00
98.0
99.0
99.7
98.7
0.00
96.7
98.7
98.7
99.8

11. CONCLUSIONS
The following conclusion can be drawn from the resultsof this research:
1-The PVC is classified as a durable product that is notaffected by humus, corrosive, acidic solution, basic
solutions, saline solution, solvents and chemicals, also
is resistant to the aggression produced by theenvironmental.
2-Addition of PVC to the granular soils at different rates
leads to a decrease in soil permeability significantly andthis serves us in improving the soil layers that areplaced under buildings and roads as sub-base or subgrade layers.
3-Addition of PVC to the clay soils at different rates leadto a substantial decrease in the permeability of the soil
and this tells us to use this type of soils in the lining of
the docks used in water treatment plants as a cheap
alternative to heavy concrete material.
4-When used percentages of PVC equal to or greater than
15%, decrease the permeability coefficient, k until to
99%.
5- Addition of PVC to gypseous soils by not less than
15% lead to a decrease in permeability and significantly
and this will reduce the melting of crystals of gypsum
surrounding soil and prevents the occurrence of thephenomenon of piping that causes the collapse of thesoil suddenly.

12. REFERENCES[1] Lamb, W. & Whitman, R. 1999. Soil Mechanics
Mexico DF Limusa. 250 pp.
[2] April, N. & Castrillon, L. 1993. Acorn stabilization
with asphalt emulsions: an expeditious alternative to
produce competent bases. Universidad Nacional de
Colombia. Faculty of Civil Engineering. Bogota' DC
Labor Grands 105 pp.
[3] Aguirre, C. & Velez, M. 2005. Granular sub basestabilization using PVC waste. . Catholic Universityof Colombia. Faculty of Civil Engineering. Bogota'
DC Labor Grands 94 pp.
[4] Amaya, C. Rodriguez, G & Valdez, J. 1978. Soil
stabilization with lime and ash Universidad Nacionalde Colombia. Faculty of Civil Engineering. Bogota'
DC Labor Grands 140 pp.
[5] Caicedo, L. & Herrera, O. 2004. Studying the
behavior of hot dense mixtures made with modified
bitumen, polyvinyl chloride waste. Catholic
University of Colombia. Faculty of Civil
Engineering. Bogota' DC Labor Grands 77 pp.
[6] Gomez, R., Moreno, C. & Reyes, H. 1983.
Stabilization materials with additives of conventional
and unconventional. Universidad Nacional de
Colombia. Bogota' DC Labor Grands 140 pp.

801

[7] Areizaga, J., Cortazar, M. & Elorsa, J. 1992.
Polymers. Madrid: Editorial Sintesis 356pp.
[8] Edgar, R., Hugo,Q., Diana, F., & Leidy, A. 2006.
Influence of the inclusion of PVC waste on the CBRof subbase granular material type. Revista
Universidad de Medellin Engineering Magazine.
Vol.5 no.9 , 1-11pp.
[9] Bowles, J. E. 1997, Foundation analysis and design.
5th ed., McGraw-Hill, New York, 46, 52-53pp.
[10] Tavenas, F., et al. 1983. The permeability of natural
soft clay. Part I: Methods of Laboratory
Measurement. " Canadian Geotechnical Journal, Vol.
20, No.4, Nov., 629-644pp.
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802

Monitoring deforestation and rangeland destruction through land use
alteration during 1955- 2002, using GIS and RS in the Merek river basin, Iran
Mohammad Gheitury*a, Mosayeb Heshmati*b, Mohammad Ahmadi* and Nik Muhamad Majid**c
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*

Department of Watershed Management, Agriculture and Natural Research Center, Kermanshah, Iran.
Department of Forest Management, Faculty of Forestry, Universiti Putra Malaysia, 43400 UPM, Serdang,Selangor, Malaysia
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**
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a. m_ghatori50@yahoo.com
b. heshmati46@gmail.com
c. nik@putra.upm.edu.my; fax: +603-89432514
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ABSTRACTLand use changing is considered by decision makers and can be precise evaluated by GIS and RS techniques.
However, land use alteration should be evaluated for monitoring and curtailing the land degradation, especiallydeforestation and destruction of rangeland. This research carried out in the Merek river basin, west Iran forevaluation of land use change during 1955- 2002 using aerial photos (1955) and satellite images (TM 1989, ETM
2002). The border and area of agriculture, forest and rangeland were distinguished and interpreted for each period(1955, 1989 and 2002). Result showed that forest and rangeland suffered from accelerated destruction during 1955–
2002 period. The area reduction rate in the forest and rangeland were 64 and 172 ha/yr, respectively. In contrast, thearea of rain-fed agriculture and mixed land use (forest- rangeland and rain-fed) were increased annually 500 ha. Rilland gully erosion features usually were obviously found in converted area, especially in the rain-fed lands, indicating
soil erosion, siltation and other environmental problems such as deforestation and carbon dioxide emission. In
addition irrigated lands were increased 40 ha annually due to enhancing ground water extraction through water welldrilling. Currently, excessive water extraction cause negative balance of ground water table and consequently water
scarcity in this basin. The mix land-use was found more vulnerable to soil erosion and deforestation problems.
Keywords:
Land use Change, Merk river basin, Rangeland Destruction, SatelliteImage

1. Introduction
Land use change is become more concern for relevant expert and decision makers as the main environmental
problem around the world and has been evaluated in many countries. It is carried out using better and more accurate
techniques such as Remote Sensing (RS) and Geographic Information System (GIS). Remote sensing through aerial
photos and satellite images can be used for land use monitoring over time. These techniques are used extensively anddespite their short history they are increasing at work in modern research on land resources which without them,some environmental issues such as land degradation, deforestation and agricultural activities is more difficult and803

maybe impossible. Aerial photo was prepared for Iran in 1955 at the 1:55,000 scale and was used for evaluation offorest, rangeland and soil erosion for first time by Faculty of Natural Resources, Tehran University, Iran [2]. The
satellite images is common tool for evaluation of forest and rangeland vegetation in Iran and reduction in naturalvegetation within watershed in the
Kabir Kouh of Ilam province (Zagros
area) was evaluated using satellite
images (1989- 2002). The results
indicated that forest area was reduced
from 108,000 ha (1989) to 98,000 ha(2002) [1].
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Nowadays, experts and managers areused GIS and RS with developed
software as the common andnecessary tools for better and moreaccurate planning in how to use land.
Because these techniques can enable
us to determine with much easespecify the environmental planning.
These techniques are also valuable
tool for soil erosion study in Iran [3].
An investigation by Srivastava andGupta [5] showed that during 19942000 about 8 km2 of natural resourcesand 2 km2 of arable lands wereconverted to urban area [5]. and usechange in Catmando valley (Nepal),
was estimated during 1989- 2005
using satellite images. The result
showed 14 % reduction inagricultural lands and adversely 17%
increase in urban and rural area [4].
The same investigation in North Iran(Roud Sary) showed 7 % reduction inforest area during 1967-1994 thatmain forest clearance was foundwhere slope was less than 20% [3].
This research carried out in the Merek
sub-basin, west Iran for evaluation ofland use change during 1955, 1989
and 2002, using aerial photos (1955) and satellite images (TM 1989, ETM 2002) and the border and area ofagriculture, forest and rangeland were distinguished and interpreted for each period.
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2. Materials and methods
2.1. The study area

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This study was conducted at the Merek river sub-basin, located in Kermanshah, Iran. It is the upper catchment of the
Karkheh River Basin (KRB) within the Zagros Mountain Chains (western Iran) (Fig. 1). This watershed has an areaof 149330 ha lies between 34º 00' 38" and 34º 09' 3" N and 47º 04' 25" and 47º 22' 18" E, comprising plains, hilly
and mountainous areas with forest, rangeland and agricultural lands. The mean annual precipitation and temperatureare 481 mm and 17.7 °C, respectively. The minimum and maximum altitude above the sea level is 1360 and 2774 m,
respectively.
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2.2. Main land use
Land use alteration during 1955-2002 was evaluated for main land use including forest, rangeland, rain-fed andirrigated agriculture. The maps of land use prepared for three periods of 1955, 1989 and 2002 using aerial photo(1955) and satellite image (TM 1989 and ETM 2002). The borders of each land use verified in the field and digitizedusing GIS (Arc.Gis 9.1) and change in area of them during these three periods were mapped (Figure2). Land use on
satellite images was separated from each other using vegetation indexes (NDVI).
3. Results and Discussion
3.1. Change in the Forest Area
This investigation showed that the forest has been reduced from 13928.4 ha (1955) to 9384.40 ha (2002) which isannually 63 ha. The respective area of forest in 1955, 1989 and 2002 is 13923.5, 10135.2 and 9384.4 ha (Table 1).
This deforestation was occurred mainly due to population growth, increase in agricultural machinery and ineffective
official control as well as enhancing the land value.
3.2. Rangeland Changing
As shown in Table 1, the respective change in rangeland in 1955, 1989 and 2002, is 22031.0, 17685.0 and 13907.0
ha. However, this study revealed that rangelands were reduced annually 173 during 47 years (Table 1). Field survey
showed that the some parts of rangeland which were converted to steep rain-fed agriculture were observed in vicinityof agricultural land with slope of 10-25 present. In this process, farmers change borders between their land andrangeland and convert some parts of rangelands to rain-fed areas annually through tillage practice. This area issubjected to unsuitable agricultural activities (mainly up-down the slope plowing) and consequently siltation
phenomenon (Fig.1).
Table1. Change in land use area during three periods 1955, 1989 and 2002 in the Merek river basin
1955
1989
2002
Land use
(ha)
(%)
(ha)
(%)
(ha)

(%)

ForestRangeland
Rain-fed agriculture
Irrigation agriculture
Mix land use
Sum

6.3
9.3
62.3
2.3
19.8
100

13928.4
22031.6
94383.61520
17467.2
149330.8

9.3
14.8
63.2
1.0
11.7
100

10135.2
17685
102081.2
1887.3
17542.1
149330.8

6.8
11.8
68.4
1.2
11.8
100

9384.413907
97480.4
3383.925175.1
149330.8

3.3. Change in the Rain-fed Land
As shown in Table 1and Fig.2, the respective change in rain-fed agriculture in 1955, 1989 and 2002, was 94383.6,
102081.2 and 97480.4 ha. However, the rate of change for rain-fed lands was 66 ha/yr since 1955 until 2002 (47years). Field verification explored that the rangeland and forest area were converted to steep rain-fed agriculture
which are characterized by steep slope, low soil quality, high soil erosion hazard and consequently low crop
productivity potential. The area of rain-fed land has been increased from 94383 (1955) to 102081 (1989) andadversely faced a reduction during last period (1989-2002) due to development of irrigated area which has been
occurred among rain-fed lands (Fig.2).
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3.4. Change in the Irrigated land
The respective change in irrigated land in the 1955, 1989 and 2002, is 1520.0, 1887.3 and 3383.9 ha indicating 40
ha/yr increase in this land-use type (Table 1). Field survey showed that the rain-fed area was changed to irrigate
lands in plain areas where slope is mainly less than 5 present due to groundwater extraction. It is carried out through
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drilled wells and has been increased during recent years affecting reduction of aquifer water and chemical fertilizer
application for irrigated crops such as corn and sugar beet.
3.5. Mix Land-use
In some area, the border of rangeland,
forest and agriculture on the satellite
image was mixed and separation theirborders were difficult because of small
areas. So these pieces were selected as themix land-use. These lands occupy 25177.1
ha (19.8%) of the study area. This type ofland-use has been increased from 17467.2
ha (1955) to 25175.1 ha (2002) (Table 1).
Field survey showed that the mix land-use
is more related to land degradation,
especially deforestation over grazing andimproper plowing. In addition, forest andrangeland in these areas are fragmented
pieces and mainly are surrounded byarable lands.
5. Conclusions
The result of this study showed thatconverting the forest and rangeland toarable lands is the certain land use
alteration within the Merek river basin.
Periodic comparing the land use alteration
during 1955-2002 indicates 5544 and8124.6 ha reduction in forest andrangeland areas resulting in developmentof arable lands. During 1955-2002
13668.6 ha (9.1%) of natural resources
was degraded trough forest clearance andrangeland destruction. In addition, forest
and rangeland in vicinity of agriculturallands are more vulnerable to alterationand consequently as the critical areas forsustain environment due improper
agricultural activities and soil erosion
hazards. Furthermore, increase in irrigatedareas through water well drilling is also
related to land use change, especially rainfed area. This land use change not only
contribute to soil erosion and siltation
problems, but negatively also cause plantbiodiversity issue mainly destruction ofsome valuable vegetation species such as
Quercus sp, Agropirom sp and Festuca sp.

806

Acknowledgements
The authors would like to acknowledge Soil conservation and Watershed Management Institute of Iran, Agricultural
Research and Education Organization of Iran (AREO) and Department of Watershed Management, Agriculture andNatural Resources Research Center of Kermanshah, Iran for their financial and technical supports.
References[1] Barkhordari, J., Zarae Mehrjerdi, M., Khosroshahi, M., 2005. Trend monitoring of the landuse changes inprotected area of Kabir Kouh with use of RS and GIS (Ilam Province-Case Study). Journal of water and soil
conservation. N 2.,[2] Heshmati M and Gheitury M., 2001. Mapping erosion forms in Merek sub basin. Soil Conservation andWatershed Management Research Center.
[3] Jafarian jeloudar, Z. , Garaei, P., Kelarestaghi, A., 2006.Landuse changes detection and spatial distribution withusing GIS techniques, case study Lajim Rood Drainage Basin, Iran. Map Asia 2006 conference.
[4] Thapa, R B and Murayama, Y., 2006. Land use change analysis using remote sensing and GIS: A case study ofKathmandu metropolitan, Nepal.pp22.
[5] Srivastava,S.k and Gupta.D.2003. Monitoring of changes in land use/land cover using multi – sensor satellite
data. Map India conference.

807

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

The comparison of phytoremediation and electrokinetic methods in
remediation of petroleum hydrocarbons contaminated soil
Gholami Meade, Yousefi Kebria Daryoush
Babol noushirvani technology university, Iran
ABSTRACT
Total petroleum hydrocarbons (TPHs) are toxic micro pollutants which are resistant to environmental degradation haveadverse effect on human’s health. Multiple methods that affect different aspects of contaminant removal can improve
remediation of petroleum hydrocarbons from soils. Phytoremediation and electrokinetic are innovative methods for
degradation different contaminant in soils. Phytoremediation is insitu method that use of plants and their associatedmicroorganisms to degrade, contain or render harmless contaminants in soils or groundwater. Electrokinetc remediation
is a technology that applies current electric field in soil to generate a voltage gradient drives soluble pollutants out ofsoil by electromigration, electroosmosis and electrophoresis. In this study, the ability of phytoremediation andelectrokinetic methods in remediation of hydrocarbons contaminated soils was investigated. According to this study,
plants with wide fiber roots, enhance degradation of petroleum contaminants, and are low-cost effective method, butneed a lot of time, while electrokinetic is more effective method in degradation of petroleum contaminant in short timeand more expensive method. In short sentence, combined electrokinetic and phytoremediation technology can uptake
more contaminant, and electro-phytoremediation is the best technology for remediation of contaminated soils.
Keywords: Phytoremediation, Electrokinetic, Soils, Petroleum Hydrocarbons

1. INTRODUCTION
Over the past years, Rapid increase in population Increase
fuel consumption, industrial chemicals, fertilizers andpesticides. Despite the use of these materials, part of thesematerials into the environment, and Actual and potential
environmental hazards are created. One of the important
issues that different communities are facing is Soilcontamination in the environment. Soil is considered uniqueenvironment for life forms, especially plants. Soil is a natural
purified, and These features are result in physical, chemicaland biological properties, but it is long that oil and its
derivatives in transport or storage are cause the soil
contamination. Petroleum hydrocarbons are one of a varietyof environmentally persistent organic pollutants, which arehazardous for living organisms. Sources of petroleumhydrocarbons in soil are crude oil extraction, transportation,refining and fuel. Soil remediation of petroleumhydrocarbons is very costly and time consuming. Improved
purification methods, Elimination or reduction of total
petroleum hydrocarbons in soil, may require mechanicalways to stimulate microbial activity. Expansion of soils
contaminated with petroleum hydrocarbons, the effective andeconomic method is one of the environmental problems.
There are various physical, chemical and biological methods
to deal with oil pollution of soils in which many of them due
to high cost and harmful side effects are used less. Amongthese methods, phytoremediation and electrokinetic toremove oil contaminants, are more economical and yet isharmless. According to the importance of soil contamination,in this study, phytoremediation and electrokinetic in refining

contaminated soils is introduced and Performance,
advantages and disadvantages of each of the methods andresults of various researchers in this field were compared.
2. Phytoremediation
Phytoremediation is a relatively new technology andeconomic purification of contaminated soils that use Suitable
plants to remove or reduce the concentration of inorganic
pollutants, radioactive and organic compounds, especially oilfrom the environment. This method the first time in 1983 hasbeen used to remove metals and other pollutants in soil.
Phytoremediation technology is based on combining its
operations with plants and microbial community thatanalysis, transmission, and living off the pollutants
compound of soil and groundwater[1]. In Phytoremediation
technology, plant roots have stimulatory effect on microbialactivity which causes suitable environmental conditions for
growth and proliferation of the microbial population. The
result is soil degradation and further reduces emissions oilfrom soil. Plants can also supply carbon, nutrients, and ... inthe soil. Thus heterotrophic bacteria grow and decompose
pollutants[2]. In general, plants provide to purify
contaminated soils direct and indirect state. Directrefinement, plants through the absorption of pollutants andaccumulation within their cells and then degradation andanalyzed them, in result these pollutants are reduced oreliminated. As well as in indirect refining, Plants throughtheir discharge stimulate microorganisms in the rhizosphereand increase microbial degradation of organic pollutants.
Root discharge such as amino acids, carbohydrates, nucleic
acids, growth factors, enzymes and other compounds with
supply of energy resources, carbon and nitrogen requirements808

of dominant microorganism in the soil, cause accelerate andintensify the destruction of these pollutants in soil[3].
2.1 Phytoremediation mechanism
Phytoremediation on the type of substance that are refined,(metals or chemicals) are classified 7 types of stem:
phytoextraction, Phytovolatalization, Phytodegradation,
Rhizodegradation, Rhizofiltration, Phytostabilization andHydraulic control. phytoextraction is absorbed pollutantsby roots and accumulation them in the shoot andthen The plant biomass is harvested and destroyed[4].
Phytovolatalizationisthe plant uptake of pollutants and evaporation from the foliage.
Phytovolatalizationisincluding pollution absorption by plants and thenrelease
the
steam
of
pollutantsout.
The
resulting product, the vapor should be less toxic than theoriginal contaminants. Phytodegradation is performed
by uptake and metabolism of contaminants intheroots,
stems
orleaves.
This
methodis degradation Pollutants a process in which
Includingabsorption,
metabolism
andanalysisofcontaminantsinsoil,
Sediment,
sludge,and groundwater or surface by produced and released
enzymes by the plant[5]. Rhizodegradation is performed byroot absorption of pollutants and no transfer take place toshoot. Rhizofiltration is extraction of pollutants by root
from surface water, waste water or ground water bysedimentation on the root or absorbed into the roots.
Phytostabilization is due to inactivity of Contaminantmaterials in the soil. Phytostabilization is the use of green
plant for the cleanup of existing pollution in soil
through improved physical, chemical and biological soil
conditions. Hydraulic control is the use of green
plant to impress the groundwater and soil water from the wayof uptake and consumption of large volumes of water.

2.2 Factors in phytoremediation
In order to be successful and achieve the greatest reduction,in order to be successful and achieve the greatest reductionofthesecompounds
concentrationsinsoil,
various environmental factors are important. Type ofsoil and organic matter present in it, can be effective inthe biological availability of petroleum product. Soil moistureis effective on plant grows and microbial activity of plants.
2.3 Advantages of Phytoremediation

A significant advantage of phytoremediation is that a varietyof organic and inorganic compounds are amenable to thephytoremediation process. Phytoremediation can be usedeither as an in situ or ex situ application. Another advantageof phytoremediation is that it is a green technology and whenproperly implemented is both environmentally friendly andaesthetically pleasing to the public.

2.4 Disadvantages and Limitations of Phytoremediation
In contrast to its many positive aspects, phytoremediation
does have a few disadvantages and limitations. It is restricted
to the rooting depth of premeditative plants. Remediation
with plants is a lengthy process, thus it may take several years
or longer to clean up a hazardous waste site, and thecontamination may still not be fully remediated.
2.5 Plant Selection Criteria
Plants should be selected according to the needs of theapplication, the contaminants of concern and their potential tothrive on contaminated soil. Design requirements should
include the use of native plants to avoid introduction ofinvasive species. The main aim is to ensure that roots expand
throughout the entire contaminated zone. Plant tolerance tohigh contaminant concentrations is also a very importantfactor to keep in mind. The phytotoxicity of petroleumhydrocarbons is a function of the specific contaminantcomposition, its concentration, and the plant species used[6].
3 ELECTROKINETICS
The term "electrokinetics" (EK) refers to the introduction ofan electrical gradient (as opposed to a hydraulic or pressure
gradient) in the soil to mobilize or promote the migration ofwater and/or various chemical species towards the preferred
electrode. Electrokinetics as a soil remediation technology isrelatively young, having become an alternative procedure forthe removal of toxic chemical species in ionic form in the soilin the late 1980's[7]. Electrokinetics is a process that
separates and extracts heavy metals, radionuclide and organic
contaminants from saturated or unsaturated soils, sludge, andsediments. A low intensity direct current is applied acrosselectrode pairs that have been implanted in the ground oneach side of the contaminated soil mass. The electrical current
causes electro osmosis and ion migration (electromigration)and electrophoresis, which move the aqueous phase
contaminants in the subsurface from one electrode to theother. Contaminants in the aqueous phase or contaminants
desorbed from the soil surface are transported towards
respective electrodes depending on their charge. The
contaminants may then be extracted to a recovery system ordeposited at the electrode. Surfactants and complexing agentscan be used to increase solubility and assist in the movementof the contaminant.
3.1 Process and mechanisms
When an electric field is created across a soil volume, it
provides a driving force that may induce mass movement ofparticles, similar to the effect of other driving forces, such aspressure gradient, concentration gradient and thermal
gradient. In particular, the application of an electric field
causes the following main transport phenomena in soils:
Electroosmosis, electro migration, electrophoresis. All theseelectrokinetic phenomena are highly influenced by the809

surface charge densities of the soil particles, and therefore by
the soil mineralogical
Composition.
3.1.1 Electroosmosis
Electroosmosis is a bulk transport of water, which flows
through the soil as a result of the applied electrical field. The
fluid migration occurs mostly from the anode to the cathode,due to the predominance of a negative charge on the soil
particle surfaces. In fact, the electroosmotic flow is caused by
the fact that when an electric field is applied to a soil, theexcess of cations close to soil particles surface (double layer)
tend to move towards the cathode. The movement of theseions and of the water molecules associated with these species(hydration shells), imparts a net strain on the pore fluid
surrounding the hydrations shell. This strain is transformed
into a shear force because of the viscosity of the pore fluid. Insum, as there is usually an excess of cations close to soil
particles, the electric fields leads to a net force towards thecathode which results in a pore fluid flux in this direction.
Hence, the electric field causes the pore fluid to flow from theanode compartment to the cathode, producing a flux andforcing the water table to arise in the cathodecompartment[8].
3.1.2 Electromigration
The second transport mechanism generated by the voltage
gradient is electro migration, which is the Movement of ionsin the pore solution under the influence of an electric field.
Positive ions (cations) migrate towards the cathode while
negative ions (anions) are transported towards the anode.
Because of electro migration, ions tend to concentrate near
the opposite charged electrode. The electro migration of
cations and anions towards the electrode opposite in charge is
proportional to the ion concentration in the pore water
solution and to the electric field strength. The ionic mobility
is a term used to describe the rate of migration of a specificion species under a unit electric field. In soils, the rate of ionicmigration can be better defined by the effective ionic
mobility, which also accounts for soil porosity and tortuosity,
which can significantly affect ion migration[9].
3.1.3 Electrophoresis
Electrophoresis consists of the movement of charged particles
and colloids under the influence of an electrical field. When adirect current (DC) electric field is applied across a colloidal
suspension, charged particles and colloids that are suspended
in the pore fluid are electrostatically attracted to one of theelectrodes and repelled from the other. Similarly to the electro
migration process, positively charged particles tend to movetowards the cathode and negatively charged particles tend tomove towards the anode. For example, negatively charged
clay particles tend to move towards anode. Usually, forenvironmental applications, electrophoresis is less important
than electroosmosis and electromigration in terms of massflux although in some cases electrophoresis may play a role indecontamination, e.g. if the migrating colloids have thecontaminants adsorbed on them[9].

3.2 Factors affecting electrokinetic technology
Electromigration rates in the subsurface depend upon grainsize, ionic mobility, contamination concentration, total ionicconcentration, and significantly upon the soil pore water
current density and pH. The direction and quantity of thecontaminant movement is influenced by the contaminant
Concentration (anions versus cations), soil type and structure,pH, interfacial chemistry, and current density of the soil porewater. The efficiency of extraction relies upon several factors
Such as the type of species, their solubility in the specific soil,
their electrical charge, their concentration relative to otherspecies, their location and form in the soil, and availability oforganic matter in the soil. Electrokinetics is applicable inzones of low hydraulic conductivity, particularly with highclay content.
4 CONCLUSIONS
Phytoremediation involves the use of living plants and theirassociated microorganisms to remove, degrade, or sequester
inorganic and organic pollutants from soil, sediment, andgroundwater. The main advantages of this method are low
cost and ecological friendliness. However, this method islimited to shallow depths (limited by the root depth) slow
plant growth, and also solubility and availability of thepollutant. Coupling electrokinetics with phytoremediation is
aimed at increasing the availability of the contaminants andalso facilitating their transport toward the root zone. The
effects of electric field on soil pH, availability of nutrients,and so on may also help plant growth. Electrodes are placed
strategically and a low direct current or voltage gradient isapplied and the contaminants are transported byelectromigration and/or electroosmosis processes toward theplant root zone. Electrode solutions of reduced toxicity
toward plants can be used to enhance solubilization of thecontaminants. Small - scale experiments showed that theplants are not affected by the exposed electric fields and that
overall contaminant removal efficiency is controlled bydifferent geochemical reactions. More research is needed toaddress organic contaminants and contaminant mixtures andpossible effects on soil quality and biology[10].
5 REFERENCES
[1]

1.

[2]

2.

[3]

3.

[4]

4.

[5]

5.

[6]

6.

810

Cunningham, S., et al. Phytoremediation of contaminated waterand soil. 1997: ACS Publications.
Günther, T., U. Dornberger, and W. Fritsche, Effects of ryegrasson biodegradation of hydrocarbons in soil. Chemosphere, 1996.
33(2): p. 203-215.
Minai-Tehrani, D., et al., Effect of salinity on biodegradation ofaliphatic fractions of crude oil in soil. Pak. J. Biol. Sci, 2006. 9: p.
1531-1535.
Pivetz, B.E., Ground water issue: phytoremediation ofcontaminated soil and ground water at hazardous waste sites.
2001, DTIC Document.
Sun, Y., et al., Phytoremediation for co-contaminated soils of
benzo [a] pyrene (B [a] P) and heavy metals using ornamental
plant Tagetes patula. Journal of Hazardous Materials, 2011.
Gleba, D., et al., Use of plant roots for phytoremediation andmolecular farming. Proceedings of the National Academy of
Sciences, 1999. 96(11): p. 5973.

[7]

7.

Reddy, K.R. and U.S. Parupudi, Removal of chromium, nickel
and cadmium from clays by in‐situ electrokinetic remediation.
Soil and Sediment Contamination, 1997. 6(4): p. 391-407.
[8] 8.
Lynch, R., et al., Preliminary tests of an electrokinetic barrier toprevent heavy metal pollution of soils. Electrochimica acta, 2007.
52(10): p. 3432-3440.
[9] 9.
Kim, S.S., J.H. Kim, and S.J. Han, Application of theelectrokinetic-Fenton process for the remediation of kaolinite
contaminated with phenanthrene. Journal of Hazardous
Materials, 2005. 118(1): p. 121-131.
[10] 10. Shenbagavalli, S. and S. Mahimairaja, Electro kinetic
remediation of contaminated habitats. African Journal ofEnvironmental Science and Technology, 2011. 4(13): p. 930-935.
[11]

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812
USING METACOGNITIVE STRATEGIES ON LISTENING COMPREHENSION

On the Effect of Using Metacognitive Strategies on Listening Comprehension of Iranian
EFL Learners

812

813
Abstract
Listening skill in the procedure of second or foreign language learning has not carried thesame attention as the other skills. Nevertheless, it’s been apparently elucidated that as a
receptive skill, listening plays a major role in enabling the language learners to acquire
linguistic information. Relevant to this fact, the present study aimed at suggesting a way
to facilitate the process of comprehending foreign/second language listening tasks;
though so general, listening comprehension tasks in a foreign language class areperceived by the researcher so crucial and require special consideration and attention.
Accordingly, conscious application of metacognitive strategies while doing the listening
tasks is the major concern of this study, and is intended to be investigated in order toanswer the question: “Do Metacognitive Strategies have any impact on the Listening
Comprehension of Iranian Guidance school EFL learners?”
In this regard, two groups of first graders of guidance school were chosen as the control
and experimental group. The experimental group went under two months treatment of
discussing about and applying the introduced metacognitive strategies while doing
English listening tasks whereas the control group did the same listening tasks without
receiving any identical treatment. During the treatment phase, the participants got
familiar with a number of listening metacognitive strategies and were asked to apply
them in listening comprehension tasks while focusing on their difficulties. Meanwhile,
they jotted down diaries regarding their perception towards using such strategies. The ttest obtained out of the listening comprehension pre-tests and post-tests revealedconsiderable improvement in the experimental group listening comprehension
performance. The findings of this study could have implications for foreign/second

813

814
language learners and teachers as well as the designers of a foreign/second language
course books.

Key Words: Learning Strategies; Metacognitive Strategies; Listening Comprehension
Introduction
Listening skill, despite its significant role in the process of foreign language
learning, has not been often considered seriously in language teaching and learning. To
this fact, it has been called “Cinderella skill” by Nunan (1997):
Listening is the Cinderella skill in second language learning. All too often, it has
been overlooked by its elder sister: speaking. For most people, being able to claim
knowledge of a second language means being able to speak and write in that
language. (Para#1)
However, second/foreign language learners usually find listening comprehension
task such a tough activity, and difficult to manage. Ur (1984) maintains that a large
number of foreign language listeners feel anxious while involving in a listening
comprehension task because of their false expectation that they have to understand every
word they hear and puts the responsibility of alleviating this worry on the shoulders of
the teachers.
In respect to language learning in general and listening skill in particular and the
difficulties occur to the listeners, the concepts of cognition and metacognition in learning
comes into one’s mind.
Based on Miller’s (1983) point of view, cognitive process orientation lead learners
toward intellectual autonomy and its curriculum “aims to develop skills so that theindividual can examine problems, consider alternative solutions, evaluate thosealternatives, choose an alternative, then implement and evaluate the solution” (p.121).

814

815
Metacognition, on the other hand, implies the knowledge of this cognitive process thatparticularly used in the realm of learning, based on which learners are consciously got
involved in the process of learning, to the purpose of being aware of their problematic
areas and finding out appropriate ways to overcome such difficulties. (Richards &Schmidt 1985)
To this purpose, learners employ some specific strategies required for a particularsituation to help them progress, known as “learning strategies”. Among the three types oflearning strategies, metacognitive strategies are at the centre of emphasis in this study.
Metacognitive strategies play a direct role on self-directed learning and learner’sautonomy (Danuwong 2006; Shannon 2008).
As the final point, a model of metacognitive strategy training developed byVandergrift et al. (2006) is worth describing here, since it’s the primary instrument usedin the present study as a means of familiarizing the participants with such strategies.
Metacognitive Awareness Listening Questionnaire (MALQ) consists of 21 item
questionnaire which is related to five metacognitive factors: problem-solving, planningand evaluation, mental translation, personal knowledge, and directed attention. This
questionnaire helps learners raise their awareness of the process of listening
comprehension as well as the strategies they intentionally benefit from to enhancecomprehension. It can also be used for self-assessment purposes by the students ordiagnostic purposes by the teacher.
This study has been run to the aim of boosting foreign language leaners’ awarenesstowards their foreign language listening skill, and suggesting a way that could probably
help them to tackle listening comprehension tasks. The researcher states her null
hypothesis as, “There is no significant difference between the listening comprehension of
Iranian EFL learners who use Metacognitive strategies for listening comprehension andthat of those who do not.”

815

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MethodResearch Design
This study was conducted based on the quasi-experimental design in which twogroups are involved, the experimental group and the control group. The independent
variable refers to the application of a number of metacognitive strategies, and thedependant variable is the listening comprehension skill.
Participants
The intended participants of the study were 60 female students of an educational
complex who were going to start their guidance school as first graders. They weregenerally about 11 or 12 years old.
Materials
A complete sample of KET was administered to determine the participants’ English
proficiency level and to make sure that they were homogeneous in this regard. The
listening sections of two different samples of KET were employed as the pre-test andpost-test with a time interval. A questionnaire including 21- item of metacognitive
strategies called MALQ developed by Vandergrift et al. (2006) was applied as a written
source of such strategies to be used by the participants. Besides, as a peripheral purpose,
this questionnaire was used to demonstrate the frequency of the strategy used by theparticipants of the experimental group. A supplementary listening book, “Tactics oflistening” in the basic level appropriate to their level, was also applied during the practicephase.
ProcedureFirst a standard proficiency test (KET) was administered to assure the homogeneityof the participants, as a result of which, they were divided into two groups.
Then, the two groups were taken a listening comprehension test (one of the samplesof KET) as a pre-test. Besides, in order to raise their listening skill awareness, theexperimental group participants were asked to jot down self reports regarding their
816

817
listening comprehension skill and their probable weaknesses. At this step, the students
attempted to reflect on their proficiency in this skill, and many of whom came into
several problematic areas while involving in a foreign language listening task. These
reports collected by the researcher, were just considered as a warm up step without any
statistics computed on them. (Although, the reports were so much in detail and could
have been analyzed and ideally applied as the descriptive data, , the researcher just aimed
at raising their sensitivity towards listening tasks through these reports.)
As the first step of the treatment, which was considered as the introduction phase, anumber of metacognitive strategies driven from Vandergrift et al’s (2006)“Metacognitive Strategy Awareness Listening Questionnaire” was introduced to theparticipants of the experimental group, and for half of a session, each item was explained,
translated and supported by appropriate examples by the researcher; meanwhile theparticipants got involved and commented on each one.
Following this very first session, the practice phase started. During this step whichlasted for about two months, four sessions a week, the strategies were regularly discussed
and practiced during the tasks of listening comprehension. Following each sessions ofmetacognitive reflection and application while doing the listening tasks, the participants
were asked to write a diary and include their perceptions towards the use and effect of thestrategies. Simultaneously, the control group received the same listening tasks but in a
traditional way of listening activities which includes listening to the recorded instruction
and doing each section as they are required.
After this period of two months metacognitive strategy instruction and practiceduring the listening activities, the final phase began with reaching a reasonableconclusion driven from a listening comprehension test as a posttest.

Results and Discussions
The test given as a pre-test, listening section driven from a typical sample of KET led
to the data shown in the following two tables:
817

818

Table 1: results of the paired sample statistics in the pre-test

Mean

N

Std.

Std. Error

Deviation

Mean

Pair 1 E1

15.333330

3.63255.66321C1

15.700030

4.06965.74301

Note.E1 stands for Experimental group and C1 for the control group before the treatment

Table 2: Results of the T-test before the treatment
Paired Differences
95% Confidence
Mean

Pair 1

Std.

Std. Error

Interval of theDeviation

Mean

Difference

t

df

Lower

Upper

-2.52286

1.78953 -.348 29

Sig. (2-tailed)

E1 - C1 .3666

5.77440

1.05426.731

7
As it is shown, the mean difference of the two groups in pre-test, with 95%
confidence of interval, fell between the lower limit of -2.522 and the upper limit of 1.789,
indicating that the obtained results were meaningful. There must be noted that despite of
the fact that the mean difference indicated a correlation between the groups, the obtained
significance exceeded 0.05, a raised issue that will be discussed later.
Having passed the period of treatment, the experimental group as well as the control
group took a post-test which was a listening section of another sample of KET. Tables 3
and 4 demonstrate the obtained results:
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819

Table 3: Results of Paired sample statistics in the post-test

Pair 1

Std.

Std. Error

Mean

N

Deviation

Mean

E2

20.933330

2.67728.48880C2

18.866730

3.29821.60217

Table 4: Results of the T-test after the treatment
Paired Differences
95% Confidence
Std.
Std.
Mean
Pair 1 E2 - C2 2.0666
7

Error

Deviation Mean
4.57077.8345
0Interval of the
Difference
Lower

Upper

t

df

Sig. (2-tailed).359913.77342

2.477

29

.019

Accordingly, it can be inferred that with the significance of 0.01<0.05, the meandifference of the experimental and control group fell in the range of lower limit and upper
limit of 95% confidence of interval: 0.3599< m <3.7734. Thus, it can be safely claimed
that the observed mean difference was meaningful and significant.
It must be notified that as it was promised before, the computed significance in the
pre-test which exceeded 0.05 must be justified here. Based on the obtained meandifference in the pre-test and post-test, it could be concluded, as a by product, thatlacking the knowledge of metacognitive strategies while performing listening
comprehension tasks, could have a negative effect on decision making about their

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homogeneity in this skill. Actually, even a standard test seems to be incapable inreaching a logical conclusion about the homogeneity of the students.
Consequently, the findings computed from the post-test revealed a meaningful
difference between the listening comprehension of the two groups with the significanceof 0.01, comparing to the pre-test, the null hypothesis of the study which maintained that
“There is no significant difference between the listening comprehension of Iranian
Guidance school EFL learners who use Metacognitive strategies for listening
comprehension and that of those who do not”, could be rejected. This conclusion
supports the previous related studies regarding the application of metacognitive strategies
in facilitating and promoting listening comprehension of EFL learner.
The study’s implications and recommendations
The findings of the present study could lead to a variety of implications in different
fields:•It suggests syllabus designers to include different types of learning strategies
particularly metacognitive strategies along with each skill exercises, toexplicitly familiarize language learners with their required strategies andschedule several exercises to create opportunities to apply them.

•

It recommends language teachers to regularly remind learners to take benefit
from metacognitive strategies and encourage them to become autonomous
learners, who are aware of their learning processes, as well as their strengths
and weaknesses.

•

Finally language learners are suggested to reflect on metacognitive strategies
and their use while engaging in different language tasks, and seek foroccasions to apply them.

Furthermore, the researcher states some recommendations for other related studies.
Firstly, it would be more satisfying to run the same study within larger group ofparticipants, so that generalization would be safely implemented. Secondly, the number
and variety of the taught and practiced metacognitive strategies can be extended to the
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821
whole 21-items stated in the “Metacognitive Awareness Listening Questionnaire”,
obviously within more allocated time. Thirdly, this research will achieve its most
reasonable and informative results if it could be done in both quantitative and qualitative
ways of gathering data, since the learners’ individual perceptions and experiences while
getting familiar with metacognitive strategies gained via self-reports or diaries would be
as much valuable as their scores in the tests or the questionnaire, and both of which could
lead to the more comprehensive conclusion.
Finally, the researcher suggests the test makers running a research to find out more
about the idea that metacognitive strategy knowledge and use could influence a sensible
result of learners’ homogeneity through standard tests.
References
Danuwong, C. (2006). The Role of Metacognitive Strategies
in Promoting Learning English as a Foreign Language
Independently. Unpublished doctoral dissertation,
Edith Cowan University-Thailand.
Miller, J. P. (1983). The Educational Spectrum. New York:
Longman.
Nunan, D. (1997). Listening in Language Learning.
Retrieved September 15, 1997, from http://www.jaltPublications.org/tlt/files/97/sep/nunan.html
Richards, J. C., & Schmidt, R. (1985). Longman Dictionaryof Language Teaching & Applied Linguistics.
London: Longman.
Shannon, S. V. (2008). Using Metacognitive Strategies andLearning Styles to Create Self-directed Learners.
Institute for Learning Styles Journal, 1, 14-28.
Ur, P. (1996). A Course in Language Teaching: Practiceand theory. New York: Cambridge University Press.
Vandergrift, L. Goh, C. M., Mareschal, C. J. &
821

822
Tafaghodtari, M. H. (2006). The Metacognitive
Awareness Listening Questionnaire: Development andvalidation. Language learning research club, 56,
431-462.

822

Evaluation of Run-Off Supply Projects in Hamadan Province (Iran)N. Rostam Afshar1, M. Abdoli2
Associate Professor, University Malaysia Sarawak, Civil Engineering Department
Email1:anrostam@feng.unimas.my
2
Project Manager, West Regional Water Company, Hamadan-Iran1

Abstract
Water resources project is for the control or use of water. Where utilization is proposed, the first question isusually how much water is needed .This is probably the most difficult of all the design problems to answer
accurately

becauseit

involves

socialandeconomical

aspect

as

well

as

engineering.

In order to meet the various demands and of water requirements in Hamadan province which is situated inwest of Iran, more than 28 projects were defined and studied, but only few of them were adopted.
This paper presents a criteria based on the ratio of reservoir capacity to volume of earthen dam named as
feasibility factor (F) for exact evaluation of different alternatives to assure the designer, so that, to obtainsufficient quantity of water in a form which can be easily and cheaply made fit for various uses.

Keywords:
Water requirement, Reservoir capacity, volume of earthen dam, Evaluation of run-off supply, Economic
analysis

1. Introduction

such a large volume of water scarcity specially in

Hamadan province is located in the western part ofdrought periods [4].

Iran with an area of more than 19000 square
kilometers. Hamadan is a beautiful highland covered
with huge mountains and green slopes, farmlands,

In the two recent decades, in order to overcome water

pastures, snowcapped peaks , anti fertile valleys undershortages, 28 reservoir dam projects were proposed ina blue sky.

Hamadan province. Two projects completed, four

Annual precipitation is 6822 million cubic meters andprojects put under construction, 11 projects were in

4000 million cubic meters evaporate. Annual water

the feasibility studies and 11 projects rejected.

consumption in Hamadan is 2900 million cubic
meters in which ground water possess 81 percent of

2. Objectives and scope

the demand and 19 percent is supplied throughIn order to meet various demands and water

surface run-off. Reference [6] shows that, 93 percent

requirements in Hamadan province, more than 28of the total Water consumption is reserved for

projects were proposed and studied, but few of them

agricultural activities, 5 percent for urban use and 2

were adopted. There is always an essential question,

percent in industrial sector. It may be seen that there

whether these projects are economically feasible andis severe shortages of water supply through-out the

could be able to provide water shortages or not?

year. In the other words, it is necessary to provide

Therefore, the main objective of this research, is topresent a criteria for selecting those projects so that823

decision makers could be assure of water supply with

4.Analysis

maximum net benefits.

4.1

Evaluation

on

the

basis

of

physicalcharacteristics of projects
Whenever, a water-resources project is planned, it

3. Methodology
In order to provide agricultural water requirements,
few parameters such as water quantity, agricultural
pattern, irrigation method, irrigation efficiency, water
conveying systems and benefit cost ratio are to beconsidered . Usually, achievement of objective ofwater resources projects is ambiguous. In thisresearch, economic evaluation based on the existing
approaches such as supply cost and reservoir storage
planning along with suggested criteria, “physicalcharacteristics of the project”, have been used for
estimating run-off supply projects in Hamadan
Province .These methods could be used in screening
phase to assure adequate supply and economic
justification.
According to Rogers [5], the real value of water,
depends upon the user view with respect to quantityand quality and water use to which it is put. The
exact value, is the summation of intrinsic andeconomic value. In the other words, full cost of wateruse, includes investment costs, operational andmaintenance costs, a water scarcity rent and the costof negative externalities of water use [8].
For economic equilibrium, the value of water, should
equal the full cost of water. However, in practice, thevalue of water in use is typically expected to behigher than the estimated full cost. This is often
because of difficulties in estimating the externalities
in the full cost calculations. More ever, in manycases, it may be lower than full cost, since politicaland social obligations override the economic issues.
Full supply cost or average value per cubic meters ofwater which has been used in this research, includes
the cost associated with the supply of water toconsumer without consideration of externalities andalternate use of the water.

824

must be ensured that the proposed proposal is thebest, and any other possible alternative will not bebetter than purposed one. To obtain such a criteria for
selecting optimum proposal, the ratio of reservoir
capacity to volume of earth fill dam named as
feasibility factor (F=C/V) was studied and noticed
that whenever the actual ratio estimated, is equal or
greater than F,

the scheme is considered to be

successful. The average value of F for existing large
dams in Iran is “20”and”9” for Hamadan. Thisparameter could be used as a primary evaluation andproject selection. The actual ratio for two proposed
projects in Hamadan, namely “Yengjeh” and “Shirinsou” dams equaled to 8 and 30 respectively. From theabove discussion, it is clear that, Shrin sou dam isonly feasible in comparison with suggested criteria.
4.2 Evaluation on the basis of Primary investment
As already discussed earlier, the cost of one cubic
meter of water is a function of investment costs,
operational and maintenance costs. According to [1],average value (base value) per cubic meter of water
estimated on the basis of 2001's tariff given byministry of energy is equal to 520 Rials(0.065$) andin accordance with shadow price in agricultural
sector, is equal to 420 Rials(0.052$) [2].
“Reference [3], [7] show that the cost of water per
cubic meter with interest rate of 7 to 14 percent for
Yengjeh and Shirin sou dams in Hamadan isestimated to be 650(0.082$) to 1300 Rials(0.16$) and
307(0.040$) to717

Rials(0.090$)

respectively.

Comparison of the estimated water cost with thewater base value, reveals that the average cost ofwater per cubic meter for Shirin sou dams equals thebase value given by ministry of energy. Therefore, it
may be concluded that on the basis of primaryinvestment, only Shirin sou dam is economicallyproject to be uneconomical. Hence, it is necessary tofeasible.

make a very judicious estimate of the storage
capacity.

4.3 Evaluation on the basis of reservoir storage

Reservoir storage planning for Shirin sou andplanning

Yengjeh dam have been worked out and the results

The storage, capacity of a reservoir is to be planned

for Yengjeh With a volume equal to 570 thousandkeeping in view the overall objective of the water

cubic meter is shown in “tab.”1. It may be seen that

resources development. The policies and guide lines

during the months of April to October, inflowof the government in a relation to water resources(108000 cubic meter) into the reservoir goes down

development of the region are to be adhered too.

but demand for irrigation (2232000 cubic meter) is

Therefore, the determination of storage capacity

high. In the other word, out flow decreases and cannot

needs detailed and careful study as the same is depend

meet the demand. Moreover, water scarcity ranges

on a number of factors. Some of the important factors

from 26 to 100 percent throughout the year. During

are,

its

monsoon months (November to March), irrigation

distribution over time, operating policy and losses

requirements is small, and reservoir volume rises tothrough evaporation and seepage. If the planned

248 thousand cubic meter which is only 43.50 percent

storage capacity is small, it may not serve the purposeof the designed capacity of the reservoir. It may be

for which it has been designed. On the other hand,concluded that planned storage capacity of Yengje

over estimation of the storage capacity will result in

dam is small and can not meet the demand.

inflow,

desiredannualdemandandconsiderably high cost of the project rendering the“Tab.” 1- Reservoir storage planning for Yengjeh dam(Units in thousand cubic meters )MonthInflow

DemandSeptember

1120

October

16

November

Outflow

Reservoir volume

Overflow

Water scarcity
Volume%Remarks

Beginning

End

1

0

0

0

119

99

Scarcity

24

16

0

0

0

8

8

Scarcity

39

0

0

0

39

0

0

0

NormalDecember

50

0

0

39

89

0

0

0

NormalJanuary

61

0

0

89

150

0

0

0

NormalFebruary

98

0

0

150

248

0

0

0

Normal

March

145

168

168

248

225

0

0

0

Normal

April

78

408

303

225

0

0

105

26

Scarcity

May

12

672

12

0

0

0

660

98

Scarcity

June

1

480

1

0

0

0

479

99

Scarcity

825

July

0

288

0

0

0

0

288

100

Scarcity

August

0

240

0

0

0

0

240

100

Scarcity

5. Conclusions

References:

This study has been mainly directed towards[1] Abdollahi,M, Rostam Afshar, Nasser, “evaluationsuggesting a suitable criteria for evaluation of run-off

the effectiveness of runoff supply projects

supply projects in Hamadan province and the data ofin Hamdean province”, Master thesis ,Power

Yengjeh and Shirin sou dams have been used toand Water University of Technology.

present the suggested approach. It may be noted that,[2] Jafari.E, Rezvani, M, “Evaluating methods tothe existing methods (primary investment andencounter crisis in Hamadan province”,2001.

reservoir storage planning) take into account theHamedan programming and management

various aspects which are required for economicalorganization.

evaluation of different alternatives.

[3] Hamedan agricultural organization Jahad ,”As discussed earlier, the existing methods, envisage

Yengjeh reservoir dam designing report”,1996.

that, Shirin sou dams is only economically feasible.

[4] Rostam Afshar, Nasser., Fahmi, Hedayatollah,

Furthermore, in this particular case, suggested“Total predicted precipitation in Iran”,1995.

criteria(physicalwater resource research organization.

characteristics

of

the

project)

proposes that, selection of Yengjeh dam is not a
properdecisionbut,

Shirinsou

project

is[5] Rogers, p,” Water as social and Economic good”,
Global Water Partnership/Swedish

economically feasible as well as there is no waterinternational development cooperation, Sweden,

scarcity during the reservoir operation.

1998.
[6] Water affairs of Hamadan, “proper situation of

5. Acknowledgement

water in Hamedan”,2001.

The authors wish to acknowledge the faculty ofengineering university Malaysia Sarawak for theirfinancialdepartment

support,(PowerHeadof water

and

water

engineering

UniversityofTechnology, Tehran) and Managing director of wateraffairs Hamadan who collaborated to develop thisarticle.

826[7] Water affairs of Hamadan,” report of Shirin soureservoir dam design”1996.
[8] Water bureau of economic,” full supply cost ofwater”, Water management of Iran,2008.

827

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Protective measures of the monumental pine in Rikuzen-Takada withsoil-water-air-solved material coupled model
Y. Sugiyama, S. Nomura, K. Kawai and A. Iizuka
Kobe University, Japan
ABSTRACTThe coastal area of eastern Japan was hit by Tsunami due to the gigantic earthquake on March 11th 2011. Tsunami gave
serious damage to Rikuzen-Takada and drowned out famous pine woods, called “Takadanomatsubara”. But, only one
pine tree of 70,000 had survived miraculously without falling down. People decided to preserve this pine tree as themonumental pine for the symbol of recovery. However, there are some problems, groundwater rise due to land sinking
and salt concentrating from seawater. These factors could kill root system. In this study, the current situation around themonumental pine was simulated with soil/water/air coupled F. E. code applying the mass transport equation first. Next,
alternative preservation methods, pumping water up and flashing, were examined. Consequently, it was found thatpumping water up from deep inside of sheet pile is effective for inhibiting salt diffusion and flashing is effective forwashing salt away from ground. This challenge will contribute to interactions between geotechnical engineering andagronomic science.
1. INTRODUCTION
March 11th 2011, the northeast part of Honshu Island, Japan,
seriously suffered severe damage and terrible Tsunami
disasters due to the gigantic earthquake of M9.0. At coastal
area of Rikuzen-Takada, 70,000 pine trees were swept awayby the tsunami and the famous pine forest disappeared. But,
among them, only a pine tree survived and monumentally
encouraged local people. It was decided to preserve this pinetree as a symbol of recovery (Fig.1).
However, the ground surface around the monumental pine
sunk 80cm due to the earthquake and groundwater level rosenear the ground surface. There occurred concerns that roots ofthe monumental pine are soaked in seawater and salt might
accumulate around the roots resulting in letting the pine treedie. Then, the protection project of the monumental pine wasoperated. Namely, steel sheet piles (15m×15m) were driven
to prevent infiltration of seawater and pumping up the water
near the pine tree was operated to lower the ground waterlevel around the pine tree. However, these efforts did noteffectively work and it was not possible to stop that the pinetree was dying.

In this study, such measures to preserve the monumental pinetree are discussed based on a series of numerical simulations
suing the soil/water/air coupled F. E. technique. And otherpossible measures are examined.
2. Governing equations for soil-water-air-solved material
coupled problem[2]In order to simulate environmental and mechanical responsethat the monumental pine tree experiences after theearthquake, soil/water/air 3 phase mixture coupledformulation considering advection-diffusion is employed asfollows, for details, see Nomura et al.
(1)
Equilibrium ij , j 0,=σ=σ ij σ ji
Effective stressσ ij′ =σ ijN + p sδ ij , σ ijN =σ ji − p aδ ij
Constitutive equation =σ ′ D ε − c s Sijijkl kl

ij

e

1
2

(2)
(3)

Compatibility equationεkl =− (ui , j + u j ,i )(4)Darcy’s law (water)vwi =− K wij h,i =−krw kwij h,i(5)Darcy’s law (air)vai =− K aij ha, j =−kra kaij ha, j(6)

Fick’s law (solved material)

J i = − Dij c, j

(7)Continuity equation (water) v = ε S − nS − nS γ
w
v r
r
r

c (8)
1+ γ cContinuity equation (air) v =(1 − S )ε + nS − n(1 − Sr ) p a (9)
ai ,i
r
v
r
Ka
i ,i

Advection-diffusion equation− S r cεv + nSr c + nS r c + cvwi ,i + nS r J i ,i =0 (10)

Here, σ ij is total stress tensor, σ ij′ is effective stress tensor,
Fig.1

Takadanomatsubara before earthquake[1]and the monumental pine

σ ijN is net stress tensor, ps is suction stress, pa is pore airpressure, Dijkl is elastic-stiffness tensor, ε ij is strain tensor,
828

cijs is coefficient tensor, Se is effective saturation, Sr isdegree of saturation, ui is displacement tensor, ε v is volume
strain, n is porosity, γ is saturated solubility, c isconcentration, K w is unsaturated permeability coefficientij

tensor, krw is specific permeability,

kwij is saturated

permeability coefficient tensor, m is Mualem’s coefficient,

K a is compressibility of pore air, K aij is unsaturated
extinction coefficient tensor, kra is permeability coefficient
ratio, ka is saturated extinction coefficient tensor, v f andijv a are velocity of water flow and air flow, J i is density flux,

Dij is coefficient of diffusion tensor, respectively, h is total
water head, and pa is atmospheric pressure.
3. Simulations of advective diffusion
3.1 Analysis condition
F.E. mesh and geometric boundary conditions employed in
the computation are summarized in Fig.2. Analysis domain isassumed to be a half space because of geometric symmetry.
The ground water level is set to be 1.0m deep from the groundsurface as a water head boundary. Right hand side of domain
is assumed to be permeable. Climate conditions such as
precipitation and evaporation employed in the computation as
input data are indicated in Fig.3. The amount of evaporation
is estimated using Penman model[3]. But actual amount of
evaporation from the ground surface varies depending on the
degree of saturation of the surface layer. Sheet piles are
installed 5.0m deep. And it is initially assumed that seawater
under the bottom of the sheet piles contains salt of which
initial concentration is C=0.1 but the ground water above thebottom of sheet piles is fresh water containing no salt. In
order to account for bio-mechanical response of the pine treesuch as transpiration, root elements[4] are employed to model

the root system. Since the monumental pine is categorized to
be Akamatsu, transpiration ability can be estimated to be
0.01m3/day and the limit of salinity tolerance can be set to be
C=0.01[5].
The ground consists of sandy layer (0 to 6m deep) and silt
layer (6 to 14.0m deep), according to field survey[6].
Coefficients of permeability for both layers are estimatedusing Creager method[7]. The moisture characteristic curvesof both soil materials are assumed as in Fig.4. Input
parameters used in the computation are tabulated in Table.1.
Computed cases are summarized in Table.2. In CASE-1,
influence of the weather condition and transpiration is
examined. In CASE-2, the actual measure employed toprotect the monumental pine by pumping up the water from2.0m deep is simulated. In CASE-3 and CASE-4, differencesof the depth from which water is pumped up, that is, 5.0m
deep and 8.0m deep, are examined. In CASE-5, the effect of
flashing on the ground surface is examined.
3.2 Influence of weather condition and transpiration(CASE-1)
Fig.5 shows the salt concentration distribution which wascomputed in CASE-1. The white color in Fig.5 indicates thelimit value of salt tolerance of the pine. According to thecomputed results of CASE-1, the salt concentration rises
upward to the ground surface gradually. The first six months
were rich in rain comparing with evaporation but after thedifference between precipitation and evaporation becomes
smaller. The distribution of total water head is shown in Fig.6,
in which total water heads under the roots of pine and outsideof sheet pile are compared. Hydraulic gradient varies withseasons and pore water migrates in accordance with it. Also,
since the total water head at the inside of the sheet pile islower than that at the outside of the sheet pile due to the water
absorption by the root of pine, the ground water migration
occurs from the outside of the sheet pile resulting in
supplying the saline water to the inside of the sheet pile. As to
the influence of climate conditions such as precipitation andevaporation at this site, the precipitation is generally more
dominant than the evaporation and it urges the downward
flow of the ground water. However, it is not enough toprevent the migration of salt content to the area of roots ofpine tree. According to the computed results, it is predictedthat the salt concentration exceeds the limit value of salt
tolerance of the pine tree at 6 months later and the pine tree
200.0
180.0
160.0

Precipitation

140.0

Real evaporation volume

120.0
100.0
80.0
60.0
40.0
20.0
0.0
Jan

Fig.2 Analysis mesh and geometric boundary condition
829

Feb Mar Apr May Jun

Jul

Aug Sep

Oct Nov Dec

Fig.3 Climate conditions

sand
silt
sheet
-pile

Table. 1

Parameters for analysis

λ

κ

M

ν

ei

0.30

0.06

1.38

0.30

1.0

m

D(m 2 day)
2.3×10-2
2.4×10-3

k f (m day)
sand
silt
sheet
-pile

ka (m day)02

2.3×10
2.4×10-1

2.3×10
2.4×10

1.0×10-10

1.0×10-100.8(a) initial condition

1.0×10-10m : Mualem[9]Degree of saturation

1

W
Sr ettia = ng
0.3 :
A=
--9
.9,
B=
2.7
,

0.8

0.6

Dry
Sr a= i ng: A=
0.22 -16.
5, B
=7.0
,
Wet
0.2 S ti ng:
r a =0.22 A=-8.2
, B=
4.3,

Logistic curve Eq. 11)
Srf - SraSr =+S
1+exp( A+Blns) raD
Sr ryi na=
0.3 g: A
=29
.3,

0.4

0
0

50

(b) 6 months after

B=
6.7
,

Silt
Sand

(c) 1 year afterFig.5 Salt concentration (CASE-1)

100

Suction (kPa)

Moisture characteristic curves of the soil material
Table.2

CASE1
CASE2
CASE3
CASE4
CASE5

0

CASES of analysis

Weather
condition(rain-evaporation)transpiration

○
○
○
○
○

○
○
○
○
○

Depth ofpumping
water
up(m)
×
2.0m
5.0m
8.0m
8.0m

Depth(m)Fig.4

flashing
×
×
×
×
○

-5
x= 0.0mx=12.0m
3 months afer
6 months afer
9 months afer
12months afer

-10

-15

-1.01

-1.00

-0.99

-0.98

Total water head(m)Fig.6 Distribution of total water head (CASE-1)

will die. It is found that the measure of sheet pile installationand pumping up water from the inside of the sheet pile is not
sufficiently effective to protect the monumental pine tree.
Furthermore the ground water level cannot be kept lower at
the inside of the sheet pile as shown in Fig.7.
4. Simulation of pumping water up

Fig.7 Distribution of degree of saturation aroundthe pine’s root system (CASE-1)

In this section, it is examined how pumping up the groundwater influences the migration of salt content in the ground.
4.1 Analysis condition
Two pumps were installed at the location of 1.5m inside fromthe sheet pile to take up water from 2.0m deep and operated
for about 4 months from Jun 13th to Oct 12th, 2011. The
ground water table at the inside of sheet pile had been lowered
about 20.0cm during that period. Therefore, the amount of 5.0
m3 is roughly estimated to be pumped up a day. In thecomputation, it is assumed that pumping up is continuously
operated after Oct 12th in order to examine its long term
effectiveness. Also, their intake location in the ground isvaried to be 2.0m deep (CASE-2), 5.0m deep (CASE-3) and
8.0m deep (CASE-4) in the computation.
4.2 Simulation results and discussion
4.2.1 Result of CASE-2
Actual effect of pumping up water is evaluated as shown in
Fig.8, in which changes of salt concentration with time during
pumping up water are depicted. When the computed results of
CASE-1 where the operation of pumping up water is notconsidered, are compared with those of CASE-2, it is foundthat the operation of pumping up water urges worse salinity830

development around the roots of pine tree. The saline sea
water reaches the inlet of pump in a year. Fig.9 shows thetotal water head distribution. The operation of pumping up
water develops the hydraulic gradient and draws the salinity
into the area of root system of the monumental pine tree.
However, it is recognized that the operation of pumping up
water lowers the ground water table and prevents
development of higher degree of saturation in the surface
layer as shown in Fig.10. In this sense, it can be said that it is
not effective enough to suppress root rot of the pine tree.
4.2.2 Result of CASE-3 and CASE-4
The imaginary cases that the inlet in the ground for pumping

up water is located at the 5.0m deep (CASE-3) and 8.0m deep
(CASE-4) are considered. Fig.11 shows change of salt
concentration (CASE-3). The location of inlet in the ground is
the same depth with the bottom of sheet pile in this case.
Similar to the results of CASE-2, it is found that salinity isgathered around the tip of sheet pile. Fig.12 indicates
distribution of total water head. The hydraulic gradient
generates downward flow of ground water near the groundsurface but it does not much grow because of water
absorption by the monumental pine. It is not effective enoughto suppress the upward advection and diffusion of salinity.
Also, it does not much contribute to lower the ground watertable as shown in Fig.13. On the other hand, in CASE-4 that(a) 4 months after(a) 4 months after(b) 8 months after(b) 8 months after(c) 1 year afterFig.8 Salt concentration (CASE-2)(c) 1 year afterFig.11 Salt concentration (CASE-3)

Fig.9 Distribution of Totl water head(CASE-2)

Fig.12 Distribution of Totl water head(CASE-3)

Fig.10 Distribution of degree of saturation aroundthe pine’s root system (CASE-2)

Fig.13 Distribution of degree of saturation around thepine’s root system (CASE-3)
831

water is pumped up from 8.0m deep in the ground, the effectof suppressing rise of salt concentration is more remarkably
recognized than others as shown in Fig.14 and Fig.15.
However, it is not effective enough to lower the ground watertable as seen in Fig.16.

generate downward flow of the ground water and prevent riseof salinity. Simultaneously, pumping up water from 8.0m
deep in the ground, 5.0m3/day, is also considered in thecomputation.

5.1 Analysis condition
In the computation, fresh water is supplied into the groundsurface as Neumann boundary condition. It is expected to

5.2 Simulation result and discussion
When amount of water for flashing was assumed to be
5.0m3/day, it was found that roots of the pine tree rotted out.
It is because the ground around roots of the monumental pinetree is fully saturated. Then, the amount of water for flashingis assumed to be 0.01m3/day, which is the half of the amountof water pumped up. Fig.17 shows change of salt
concentration. It is understood that rise of salinity(a) 4 months after(a) 4 months after(b) 8 months after(b) 8 months after(c) 1 year afterFig.14 Salt concentration (CASE-4)(c) 1 year afterFig.17 Salt concentration (CASE-5)

Fig.15 Distribution of Totl water head(CASE-4)

Fig.18 Distribution of Totl water head(CASE-5)

Fig.16 Distribution of degree of saturation aroundthe pine’s root system (CASE-4)

Fig.19 Distribution of degree of saturation aroundthe pine’s root system (CASE-5)

5. Simulation of flashing (CASE-5)
In this section, it is examined how fresh water flashing effects
dilution of salinity and protect the monumental pine tree.

832concentration around the roots of pine tree is well suppressed.
The hydraulic gradient does not much developed as seen in
Fig.18. Therefore, it can be said that the use of flashing andpumping up together is effective to save the monumental pinetree. In addition, it is found that the salt concentration does
not exceed the limit value of salt tolerance of the pine tree.
Fig.19 shows distribution of the degree of saturation. The
ground around roots of the monumental pine tree does notreach fully saturation in a year. This implies that the root rot
will be avoided. However, on the other hand, seamless
operation of pumping up water and flashing at the same time
would not be actually feasible from economical viewpoint.

[7] Creager, W.P., J.D. Justin and J. Hind: Chap.16 soil tests
and
their
utilization,Engineering
for
dams,
Vol.III:Earth,Rock-fill,Steel and Timber Dams, John
Wiley and Sons, pp.645-654, 1944.
[8] Kawai, K, Wang, W, Iizuka, A. The expression ofhysteresis appearing on water characteristic curves andchange of stresses in unsaturated soils, Journal ofApplied Mechanics JSCE, Vol.5, pp.777-784, 2002.
[9] Mualem, Y. A new model for predicting the hydraulicconductivity of unsaturated porous media, Water
Resources Research, 12(3), 514-522.1976.

6. Conclusions
In order to protect a monumental pine that survived fromtsunami disaster in Rikuzen-Takada, possible measures were
examined through a series of numerical simulations using
soil/water/air coupled and advection-diffusion F.E. code,
DACSAR. First, actual protective measure employed in thesite was examined. And other possible measures werediscussed. Concluding remarks are as follows.
1. The sheet pile of 5.0 m long is not sufficient to interrupt
inflow of seawater into the area of roots of the pine tree.
Some supplementary measures are required.
2. In the operation of pumping up the ground water, thelocation of inlet is important. In the case of relatively
shallow location of inlet for pumping up water, it does
not work so as to prevent migration of salinity into thearea of roots of the pine tree. It urges rise of salinity
instead. On the contrary, too deep location of it does notwork to lower the ground water table. It possibly causes
root rot of the pine tree because roots of the pine tree are
immersed.
3. The use of fresh water flashing together with pumping
water operation is effective. Continuous supply of fresh
water into the ground surface prevent rise of salinity.
And pumping water operation works so as to lower the
ground water table. However, its feasibility is stillquestionable from economical viewpoint.
7. REFERENCES[1] http://www.iwatetabi.jp/event/detail/03210/115.html[2] Nomura, S. Mathematical model of soil/water/air
coupled problem for unsaturated soils considering masstransfer within pore-water, Proc. of the 5th Asia-PacificConference on Unsaturated Soils, Vol.1, pp.287-292,
2011.
[3] Penman,H.L. Natural evapotranspiration from open
water, bare soil and grass，Proc. Roy. Soc. London,
A193,pp.120-146,1948[4] Kawai, K, Iizuka, A, Simonaga, K, Tachibana, S. The
influence of uptake of vegetation on the ground, Journalof Applied Mechanics JSCE, Vol.10, pp.375-381, 2007.
[5] Larchel,W.
Okophysiologie
derPflanzen,
Springer-Verlag Tokyo, p-350, 2004.
[6] The business report of current situation survey of theground and seawall in Rikuzen Takada, 1999.

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Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Effects of Biosurfactants in Electrokinetic Remediation ofcontaminated soils
Bour Moslem1 and Yousefi kebria Daryoush2
Babol Nushirvani University of Technology, Iran
ABSTRACTOne of the most important In-situ remediation methods of contaminated soils is electrokinetic remediation, that byusing a direct electric field with various mechanisms (electromigration, electroosmosis, electrophoresis andelectrolysis), the pollutants move in and exit from the soil. Electrokinetic remediation in soil has been limited due tolow solubility of organic materials in water and their non-ionic nature. The use of surfactants (agent active surface)
increases the remediation efficiency with increasing the solubility of insoluble liquid petroleum compounds in waterand enhances their bioavailability. Biosurfactants are produced by microorganisms and contain a hydrophobic andhydrophilic moiety. They increase the water concentration of hydrophobic compounds and mass transfer rate bydesorbing and emulsifying the adsorbed hydrocarbons to organic materials of soil. Therefore, they help that
acceleration in biodegradation be occurred. Studies show that the extraction efficiency of oil compounds by using ofsynthetic surfactant could be increased 1.5 to 5 times in comparison with water, whereas the ability of Biosurfactant
for removing of petroleum contaminants in the soil is 1.9 to 2.3 times more than synthetic surfactant.
Keywords: Electrokinetic, Remediation, Biosurfactants, Contaminated soil
1. INTRODUCTION
It’s a long period of time that the oil materials and its
derivative in transportation or storage develop pollution insoil. Oil pollution is an inevitable consequence of a rapid
increase in population and industrialization processes that
Soil contamination by substances that followed
Hydrocarbonate and vast of oil exploration and refining
facilities in around the broad and topical routes oftransmission of this material is visible. In addition to directemissions of these pollutants, dust from fuel emission withoil, over the years been able to add toxic and harmful
substances in that soils. Biotechnology is one of the
branches of the oil business and today the list of services
offered by companies active in environmental issues is seen,is the Biological clear waters and soils contaminated withpetroleum compounds. According to Iran's oil resources,
production, Refining and distributing petroleum products, it
is heavily exposed to oil contaminated soil. Based on thisand due to the destructive environmental effects, cleanup oil
contaminated areas has great importance. Mostly purifying
the contaminated areas is done in two ways including insideand outside of place. The electrokinetic process is one of themost important ways in clean up the contaminated soils,
faces with limitation because of low solubility of organics inwater and their non-ionic nature. In order to enhance thebiological availability of oil pollution and overcome to thehydrophobicity of their property, surfactants are used. The
purpose of this study is to evaluate the impact ofbiosurfactants at the remediation of oil contaminated soils
using electrokinetic method.
1.1 Soil, properties and contamination

Soil is one of the important resources and valuable nature
which is produce human diet of about 95 percent. Without
healthy soil, Life and living on earth would not be possible,
therefore having planning for a healthy soil is essential forhuman survival. Soils are Mixture of mineral and organic
which are formed of Degradation and destruction resulting
from weathering of rocks, and generally are composed oftwo sections living and dead. Typical soil has 50% minerals
and organic material and has 50% air and water that fills theempty spaces in the soil and keep living organisms in thesoil. The type and combination of the soils in various regions
is different depending on the district condition.
Soil pollution is the presence, distribution or mixing one ormore foreign substances in the soil, while the amount andtime that changes the quality of physical, chemical andbiological of soil, so that it is dangerous for humans or otherliving creatures or plants or works and buildings. In otherwords Soil pollution is any changes in the characteristics ofthe soil forming components So that it is impossible forusing. Most oil pollution of soil resources is seen aroundrefineries. Much oil materials will penetrate deeper into thesoil, more difficult to fix its pollution and its costs will be
multiplied. The reason of Retention and stability of oil
pollutants in soil properties returns due to thehydrophobicity of compound. An overall emission in the soil
takes place in two stages; Initial phase that is short andquickly terminated then began the long phase and slowly
goes away.
1.2 Soil decontamination methods
Nowadays the development of effective methods to soil
remediation is allocated an important part of the research to
the environment. Most remediation methods have been
established and protection of human, animal andenvironmental liberation from exposure risks to toxic

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substances. Many methods are used for the purification ofcontaminated soil, has resulted of research and experience inother sciences.
Broad range of technologies exists for purifying
contaminated soils, but is generally divided into two
categories:
1- In-situ Remediation: To do the process of clearing incontaminated site and2- Ex-situ Remediation: The contaminated soil is transferredfrom contaminated area to another location for remediation.
2. Electrokinetic Remediation
Electrokinetic process consists of applying an electric field
to the soil environment by placing two electrodes into thecontaminated soil that flow directly between the electrodes
is low and provide conditions that are different mechanismsof pollutants during pregnancy in the state will move toward
the electrode with opposite charge. This process has widelyused today to separate contaminants in soil and is classified
as one of the methods of physical - chemical cleanup of
contaminated soils [1]-[2]. Electrokinetic is a controlled
method which is applicable in every area so that the cleanupof contaminated soils would be possible under buildings
constructed without damage to the building [3]. This process
was first developed by the Reuss then by Helmotz, Pallat
and Smolochiwski the theory was presented. For soil andsludge samples were also used by Kasagrand [4]-[5]-[6]-[7].
2.1. Mechanisms of Electrokinetic process
Electrokinetic process has different mechanisms (showed in
Fig 1) such as:
1- Electrolysis: Chemical reactions are performed in theelectrolyte around the electrodes.
Establish the electric field with the following reactions
occurs at the electrodes during electrolysis.
In Anode:2H 2 O – 4eO 2 + 4H+
In Cathode:4H 2 O + 4e
2H 2 +4OH+
H production at the anode is decreased causes reduction in
pH. Acid front generated at the anode will migrate to thecathode during the process when water is electrolysed.
Maybe PH at the anode decreases to less than 2 and at thecathode increases to more than 12. As a result, ion of H+ will
move toward cathode and ion of OH- toward anode. Ion
speed for Ion of H+ at least is double Ion of OH- that theflow Electroosmosis becomes more. If the buffering capacityof soil is not acidic side of the transmission delay Ion of H+
transport within the sample dominates the acidic [8].
2- Electromigration: Motion of ions under the influence ofan electrical gradient.
3- Electrophoresis: Motion of charged particles under the
influence of colloidal and electrical gradient.
As for the similar mechanisms of ion migration andelectrophoresis, generally, both as a general mechanism areconsidered. In general, charged particles including colloids
and ions move toward the non-pole. Metal ions, Ammonium
ions and positively charged organic compounds move
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toward the cathode and anions and negatively charged
organic compounds move toward the anode. After certain
time, the pollutants accumulate in the Cathode and anodeand with adequate measures isolation would be available [8].
4- Electroosmosis: Water transfer between soil cavities
under the influence of an electrical gradient.
The effect of electric field applied along the soil mass,
Positive charges towards the negative pole (cathode) andnegative charges toward the positive pole (anode) areabsorbed. A number of additional positive charges on thesurface soil are pregnant. When this additional charges movetoward the negative pole, Water due to friction force withthe particle moves toward the negative pole. There is also atrend toward the positive pole, but because the water flow
toward the cathode is generally smaller and is called
electroosmosis [8].

Fig.1 Mechanisms of electrokinetic process
2.2. Format Mechanisms for various pollutants
In Electrokinetic process any pollutants are removed fromthe soil to one of the following statements:- Minerals by the mechanism of ionic migration in electricfield move toward opposite poles and are removed from thesoil.
- Polar organic compounds are removed from the soil byionic migration.
- Non-polar organic compounds soluble in water due tosolubility in water, by electroosmosis mechanism get sent
the cathodes and are removed from bulk soil.
- Non-polar organic compounds insoluble in water due to theuse of solvents and surfactants in the Andes as extraction
materials they move from the anode to the cathode go to thecathode.
The main purpose of the application process to remove
contaminants such as heavy metals Electrokinetic, anions
and polar organic compounds in soil, sludge, mud, dredged
sediments at sea etc. this method has better yields in soil
with low permeability soils that soils are generally clay or ismixed silt and clay that like, Capabilities of pollutants will
be better to move some of the mechanisms Electrokinetic.
On the one hand, Heterogeneous soils with low permeability
because of the limited movement of pollutants, Bacteria andnutrients, bacteria cannot decompose pollutants therefore theneed for special arrangements is unavoidable. The problem835is removed with direct current electricity generated by aseries of electrodes in the soil mass, because the flow isresulted in various phenomena such as: osmotic movement,
ion migration and the particle motion. For removing organic
contaminants by using the Electrokinetic method, Molecularsize and tendency to accumulation or adsorption to the soil
surface are the main factors controlling soil improvement.
The high concentration of ionic particles in electrochemical
reactions, Migration of electrolyte and water transfer
between ions and water are friction factors. Obviously,
Electrokinetic process efficiency depends on organic
removal of solubility their being polar [1]-[8].
2.3. Advantages and disadvantages of Electrokinetic
process
The main advantages of the Electrokinetic processapplication:
This method is very fast and relatively with high efficiency
and easily prevents the spread of contamination.
As well as out of place in situ is applicable.
Unlike other in-situ recovery methods that are not effective
in soils with low permeability or small seed, Electrokinetic
method due to the negative surface charge of clay particles,
has the highest efficiency in loam soil. So Electrokinetic isan effective method for moving water, and colloidal ions insoils with a small seed.
In addition to removing the decontamination, it has so muchapplication such as: compassing the contamination,
Inundation and concentrate dewatering sludge and extracting
industrial metals in mineral and environmental sectors.
The most important problems in the Electrokinetic process
are the Change in pH near the cathode and the sequestrationof metals in this range. This phenomenon begins when large
amount of ions OH- Produced at the cathode and thesurrounding medium is strongly alkaline. However,production H+ around the anode, pH is greatly reduced.
Usually in this case a sharp pH gradient is created between
two electrodes. pH changes causes disruption in the transfer
process, Biological activity of bacteria and adsorption andbiodegradation of organic matter in the soil [8].
3. Surfactants
Surfactants are compounds of amphiphilic that free energyof the system by the replacement of large molecules withhigh energy in the medium reduces. They are using for soil
washing because of the ability to move the pollutants. They
have a slight hydrophobic environment with low interest
bulk and hydrophilic groups which are willing to fly to theenvironment (showed in Fig 2).
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Currently the use of surfactants has extensively developed inindustry. In industrial applications, the ionic surfactants areoften classified based on their charge which has at neutral
pH when dissolving in water. These categories include
anionic surfactants, cationic and neutral or Amphoters.
Petrochemical are most traditional users of this material as a
facilitator for oil removal from tannery wastewater. In thisapplication, Petrochemical surfactants increase the solubilityof compounds. This material also for floating materials areused in mining and the pharmaceutical industry and health[9].
Increasing the solubility of oil and liquid petroleum
compounds which are water insoluble is the main uses of
surfactants. Increasing the solubility by reducing the surface
tension and interfacial tension in the layer between the waterand oil is carried out with petroleum compounds. Surfactantsin the intermediate level solid–liquid, liquid-liquid or
vapour-liquid are gathered. There are two distinct phases
between the interfaces hydrophobic component of water
when hydrophilic component towards a solution, willaccumulate on surface. The effect of surfactants isdetermined by their ability to reduce surface tension,
criterion for the surface free energy per unit area for thebringing the bulk phase to the surface of the molecule [9].
Solubility of non-aqueous phase liquids by surfactants with
reducing the surface tension between the joint surfaces to
air-water and water–oil is increased. There is a correlation
between surfactant concentration and surface tension. Whenthe concentration increases the surface tension is reduced toa minimum value. Surfactant concentration at this point
called as the critical micelle concentration (CMC).
Surfactant properties including increased solubility, reducing
surface tension and the low critical micelle concentration.
One of the important characteristics is the criticalconcentration cleansing that depends on many of their
physical properties such as emulsion formation, compounds
dissolve oil and petroleum, Foaming, cleaning, the surface
elasticity and between the surface elasticity. These
properties are considered important criteria to determine theappropriate surfactants in soil washing process. Because of
surfactant presence, less work is needed for moving themolecule to the surface and the surface tension decreases.
Surface tension is related to concentration of surface
activators until the concentration reaches to a critical
micelle. Effective surfactants have with the less critical
Micelle concentrations. CMC is defined as the minimum
concentration required initiating the creation of micelle.
Actually, CMC is the maximum monomer surfactant inwater that influenced by pH, temperature and ionic strength.
Rising temperature causes reduction in CMC and thusincreases the efficiency of non-ion surfactants and also
increases the solubility of ionic surfactants. Increasing thesalt concentration in water can also reduce the amount of
CMC for ionic surfactants but has no significant effect on
their non-ionic types [9]. Surfactants in terms of production
are divided into two categories: Synthetic surfactant,
Biosurfactant.
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836

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Fig.2 Schematic of the Amphiphilic molecule of surfactants

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3.1. Synthetic surfactants
Various types of cleansing which are produced using various
chemical compounds are known as synthetic Synthetic
surfactants. Synthetic surfactants are produced in three types
including anionic, non-ionic, Cations that all of which areused for soil washing. SDS, A synthetic anionic surfactant
that cleans the soil in the field is the most common
surfactant [10]. In addition, non-ionic surfactant due to theirlower CMC than the anionic surfactant., Higher degree of
surface tension reduction and relatively constant properties
in the presence of salt have better performance and less need
to use and consequently are used in improved cleaning
process with surfactant assistance. One of the non-ionicsynthetic surfactants can be named as Triton X100. Out of
the cleaners Benefits in the industry, Use of surfactants inthe environmental engineering can be count on thefollowing:- Make improvements in the toxins absorbed from the soil
- Improved desorption of organic contaminants in soil
- Desorption of organic contaminants from soil
Like many artefacts of human manufacture of synthetic
surfactants significant negative have effects on biologicalactivity Ecosystems and our environment. One of the major
disadvantages is the effect that the use of this material is in
the burning eyes and skin. Although different surfactants
have effects, the negative effects and toxicity of thesesubstances are found for living cells. All arrangements are
the use of alternatives reason in addition to preventing theirspread. One of the alternatives for these surfactants aresynthetic surfactants that have similar properties as
Biological surfactants And their negative impact is very low
because they are one of the nature part [11]-[12].
3.2. BiosurfactantsThose surfactants that are obtained from natural compounds
called biological surfactants. These surfactants besides thecleaning animals and plants, by microorganisms, yeast or
bacteria to biological methods, of different nutrients,
including sugar, oils, alkenes, and wastes are produced.
Most Biosurfactants are a kind of non-ionic or anionicsurfactant and include one or several parts of lipophilic andhydrophilic [1]-[13]-[14]. The Lipophilic part can be aprotein or a lipid with a large number of sidelonghydrophobic chains. But often the hydrocarbon chain of afatty acid with 10 to 18 carbon atoms is formed. The Water
friend sector can be an Esther, A hydroxyl, or a phosphate ora carbohydrate group and or a Hydroxilit. Biosurfactant
based on chemical structure and type of producing
microorganism have different types and are classified Ingroups of glycolipid, phospholipids, lipid complexes – Polysaccharin, Lipoprotein – Lipopidieta Neutral lipids and fatty
acids Hydroxilat [1]-14].
3.3. Work of biosurfactants in removal in oil
combination
When biosurfactants is added to the soil it collects to nameof micelle is gathered that polar section toward out andhydrophobia tails toward in is stand. In this state thehydrophobic hydrocarbons located in the core of this micelle
and in the form of a hydrophilic colloid in coming and areseparated from the soil surface if higher concentrations ofsurfactants, afar said colloids even are eventuated tosolution[15]-[16]-[17]-[18]. This is exactly the samefunction as the soap is fat food. The fat particles in the soap
micelle and with water from the soil surface are removed.
Biosurfactant to remove petroleum compounds from soil by
two main mechanisms are strengthened:
1- Mobilization: The motion of particles in the oil
2- Solubilisation: Solubility of oil components in water
Mechanism of stimulation usually occurs at concentrations
below the critical concentration and micelle surfactants.
Reduce surface tension and between the surfaces, reductionof capillary force, Moisture variability and increase thecontact angle is. Surfactant monomers increase the contact
angle of between hydrophobic pollutants in soil, theseparation of contaminants develop from soil particles andultimately are removed the oil out of the soil. In thismechanism, in concentrations below the CMC surfactants
are reduce the surface tension between surface the phases of
surface and air-water, oil-water and soil-water. Surfactants
that are in contact with the surface between two phases ofsoil and oil, Contact angle increases and are decreases thetensile force between soil and oil levels. And therefore themain cause of stick to the soil and oil particles that thecapillary force is reduced. The simulation mechanism
depends on of ionic time surfactants; therefore, the metals inthe soil to absorb anionic surfactants may reduce theireffective concentration in the soil [18]-[19]. With the captureof surfactants for their effect on soil and soil stimulation
may occur with the highest concentrations of surfactants.
When the amount of surfactant is more than criticalconcentration, pollutants from soil to cores of micellehydrophobic are removed by surfactant. Crude oil Solubility
in water with production increase fields monomericsurfactants. Hydrophobic ends of surfactant molecules toform a cluster inside the building had accumulated micelleand love water ends them are located bottom exterior
surface in contact with the aqueous phase structure if
micelle, the internal part of a micelle as a medium suitablefor hydrophobic molecules in the organic material comes
and is capable oil particles to separating from the soil
surface and into solution. This process is known as
dissolution. Biological surfactant to facilitate the desorptionof oil pollution of soil or by spreading oil particles attachedto the soil surface droplets into the liquid phase so thatmicroorganisms are easily available be modified to process
and analyze biological compounds, petroleum contaminatedsoils will help. Given the existence of different types ofhydrocarbon molecules complexity and extent of physicaland chemical properties are very low solubility in crude oiland most of them and lack of access to the very low
interfacial tension in the soil washing and effect of The
mechanism of dissolution of crude oil in oil to remove

837

contaminants from soil, Is almost negligible to minor. The
removal of crude oil composition of the soil would be
related to the mechanism of stimulation [18]-[19].
3.4. Biosurfactants benefits of in comparison withsynthetic surfactants
The tendency to Bio-surfactants in comparison withsynthetic surfactants has increased day to day. This is partly
because of better indicator of bio-surfactants in comparison
with synthetic surfactants which can be listed as below: Facilitate procurement of renewable natural resources- Less toxicity and excellent surface activity than synthetic
Surfactants- Biodegradable- Environmental sustainability of industrial effluent
discharged into the environment of the soil washing process- And ability to operate at high temperatures- Ability to maintain the high pH, more than 12- Ability to tolerate high concentration of salt dissolved inwater- Ability to remove oil contaminants from soil, 1.9 to 2.3
times more than synthetic surfactants [20]4. CONLUSION
1) Decisions about choosing the best method or methods
normally, Shall be based on cost And cannot be said withcertainty which method is more economical than other,because some of the technologies depending on the soil canbe very costly.
2) Electrokinetic process due to the possibility of being
implemented in the site and simplicity has widely used inrecent years. Also, the above process is able to remove awide range of pollutants from the soil.
3) Low solubility and non-ionic nature of the soil organic
matter in the water causes electrokinetic process to belimited.
4) Using surfactant and increasing the solubility of liquid
petroleum compounds insoluble in water and enhance theaccessibility of their biological can increase the cleanup
efficiency.
5) The main biological features of soil after clearing process
remains unchanged.
5. REFERENCE[1] O. Gozini, A. Palza, M. C. Lob, L. Di Palma,
Electrokinetic remediation of gasoil contaminated soil
enhanced by rhamnolipid, Springer Science+Business Media
B.V. 40:1239–1248, 2010.
[2] Riser-Roberts E, Remediation of gasoil contaminatedsoils: biological, physical and chemical processes, Lewis
Publisher, Boca Raton, 1998.
[3] Rosanne M., Paszkuta M., Adler P.M. (2006)
Electrokinetic phenomena in saturated compact clays, Journalof Colloid and Interface Science. 297: 353-364.

[4] AEC, 1997. In – situ Electrokinetic Remediation for
Metals Contaminated Soils FY96, Annual Report, No.SFIMAEC-ET-CR-97013, pp.87-88.
[5] Chung, H.I., Kang, B.H., 1999. Lead removal fromcontaminated marine clay by Electrokinetic soil
decontamination. Engineering Geology, 53: 139 – 150.
[6] DOE, Office of E.M., 1995. Electrokinetic Remediationof Heavy Metals and Radionuclides Technology Catalogue.
second edition, EM-0235, 201-203.
[7] EPA, 1997. Recent Development For In-situ treatment ofmetal contaminated Soils.
[8] Virkutyt, J, Sillanpa, M, Latostenmaa, P., 2002,
Electrokinetic soil remediation – critical overview, Thescience of the total Environmen, 289, 97 – 121.
[9] Rosen, M.J., Surfactants and Inter facial Phenomena,
John Wiley, New York,1989.
[10] Giannis A, Gidarakos E, Skouta A, Application of
sodium dodecyl sulphate and humic acid as surfactants onelectrokinetic remediation of cadmium-contaminated soil,
Desalination 211:249–260, 2007.
[11] Urum, K. and Pekdemir, T., Evaluation of
biosurfactants for crude oil contaminated soil washing,
Chemosphere, Vol.57,2004,pp.1139-50.
[12] Paria S, Surfactant-enhanced remediation of organic
contaminated soil and water, Advances in Colloid andInterface Science,Vol.138, pp.2458, 2008.
[13] Catherine N.M, Environmental applications forbiosurfactants, Environ. Pollut.133 183-198, 2005.
[14] Kosaric N, Biosurfactants and Their Application for
Soil Bioremediation, Biosurfactants for Soil Bioremediation,
Food Technol. Biotechnol. 39 (4): 295,304, 2001.
[15] Kuyukina M.S, Irena B. Ivshina, Sergey O. Makarov,
Litvinenko L.V., Cunningham C.J., Philp J.C, Effect ofbiosurfactants on crude oil desorption and mobilization in asoil system, Environment International. 31, 155,161, 2005.
[16] Mulligan,C.N., F.,Eftekhari, Remediation with
Surfactant Foam of PCP-Contaminated Soil, Engineering
Geology, 70 (3-4):269-27, 2003.
[17] Mulligan, C.N., R.N.,Yong, B.F.,Gibbs, Surfactant
Enhanced Remediation of Contaminated Soil: a review,
Engineering Geology 60, pp. 371-382, 2001.
[18] Mulligan, C.N., Environment Application for
Biosurfactants: Environmental Pollution, Vol. 133, pp. 18319, 2005.
[19] Rahman,K.S.M., et al, Bioremediation of Gasoline
Contaminated Soil by a Bacterial Consortium Amended withpoultry litter, Litter Coir Pith Rhamnolipid Biosurfactant,
Bioresource Technology, 81(1):25-32, 2002.
[20] Urum,K., Y.,Pekdemir, M.,Çopur, Optimum Conditions
for Washing of Crude Oil Contaminated Soil with
Biosurfactants Solutions,Trans IChemE, Vol. 81, 2003.

838

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

O2117 Change in structure of ground-beetle assemblage on river areas after
construction of biotopes using the technique, “Nature oriented river works”
Michko Masuda1, Takaaki Tsukada1 and Fumitake Nishimura2
Department of Civil Engineering, Nagoya Institute of Technology, 2Department of Environmental Engineering,
Kyoto University.

1

ABSTRACT
Nature oriented river works means the river is managed for maintenance and the creation of a native habitat of the living
things that the river originally maintains and various river spectacles. The kind of river constructions has been tried byusing the living things such as dragonflies, butterflies, birds and fishes. The assessment is dependent on an
investigator’s skill, and does not serve as quantitative data. Then we tried environment assessment using ground-beetle
assemblage. There are many kinds of species in the ground-beetles. It is suitable for prediction of the influence of an
artificial disturbance, and the index for evaluation. Some pit-fall trap were set up at some different vegetation in the twobiotopes and the structure of ground-beetle assemblage was investigated over six months. As a result of the followings
become clear, 1)ground-beetle assemblage had the same tendency as the natural river areas. 2)Ground-beetle
assemblage on artificial area is reflection the characteristic of local environment and we can use it for environment
assessment as an index.
Keywords: Environment assessment, River management, Biodiversity investigation, Environmental indicator

1. INTRODUCTION
The river law of Japan was revised in 1997. After then"Nature oriented river works" has been done as a stream
management that considers the environment. "Natureoriented river works" means the river is managed formaintenance and the creation of a native habitat of the living
thing that the river originally maintains and various river
spectacles. Disturbance, destruction and reduction of thehabitat of wildlife are take place by river improvement. In
order to evaluate the influence by an artificial disturbance tobiota, it is important to monitor change of the kind phase of an
object living thing, and to compare disturbance order. Until
now, the researches of environment assessments were carriedout using aquatic insects, butterflies and dragonflies [1]-[9].
However there are some problems in using dragonflies andbutterflies. Dragonflies seriously depend on the water quality,and butterflies flight area are very vast, then the quantitative
investigation are not easy. It is not necessarily appropriate toconsider it as the index living thing universally used for much
environment [10].
The ground-beetles in Japan are which has very many
numbers of species, and numbers of individuals and in which
ecological research is also progressing [11]-[13]. The groundbeetles are a general term for beetles which mainly walkalong earth surface, carry out surroundings ingestion and areliving, such as a ground beetle, carabit beetle, a burying
beetle. Moreover, the ground-beetles forms the community
structure according to a slight environmental change of eachhabitat, and reacts to an environmental change quickly very
sensitively [11] [14]. Then, it is indicated that thecharacteristic of the ground-beetles is suitable for predictionof the influence of an artificial disturbance and the index forevaluation [15] [16]. Research about the ground-beetles as theenvironmental indicator has been done centering on Europe
from the 1960s [17]. In Japan, it is these days that research on
the environmental indicator of the ground-beetles has been
done [18][19].
However, there are few researches on the environment
assessment using the ground-beetles in the new biotopes
developed by "Nature oriented river works". So, in thisresearch, it aimed at clarifying analysis of the community
structure of the ground-beetles in an investigation place,environment assessment, and correlativity with a plantcommunity by investigating quantitatively the number of theground-beetles of the biotope circumference of Yadagawa,river and the Shonai river that flows through northern
Nagoya, the center of Japan.
Furthermore, from a viewpoint of conservation of theground-beetles, it advocates considering it as one index of the
future production of the “Nature oriented river works”.
2. Materials and Methods
2.1 Study site
The study was carried out on the two biotopes at left bankof Yada River at Joganji, Nagoya City (35o 21’ 60’’N, 136o
91’ 70’’E, altitude 9 m). and at left bank of Shonai River at
Nishibiwajima, Kiyosu City (35o 19’ 03’’N, 136o 85’ 80’’E,
altitude 3 m) in Fig.1. Yada river is a urban river in the Shonai
River water system of the extension 23km and 115 km2 in thevalley area. Shonai river is a urban river water system of theextension 96km and 1010 km2 in the valley area. Each pointand environmental condition showed in Table 1.

839

2.3 Statistical analysis
In order to analysis of the census, five diversity index, anevenness index, two expected value and a disturbance index
were used. Each index was calculated along the Table 2.
Table 2. The equations of each index

Statistical value
Diversity
Simpson index

Equations

Shanon-Winner index
evenness
Expected
DisturbancePielou index
Chao species richness
index
Ishitani index

n i : the number of entities belonging to the ith type, N : the total number ofentities, S : the total number of species, a : the number of species observed
only one individual per species, b : the number of species observed more than
one individual per species. I i : Environmental index value of the ith type, N ij
: the number of the ith species in jth investigated plots.

Detrended correspondence analysis (DCA) was conducted
to evaluate variations in the ground-beetles assemblages
among different vegetation plots. DCA was conducted with arearrangement at each plot, and scored.

Fig. 1. The map of study site, Yadagawa biotope and Shonai
biotope.

Table 1. The environmental condition of each study plots.
Biotope site
Yadagawa

Vegetation type
Grass land
Floodplain forest
Grass land
Floodplain forest
wetland

3. Results

Disturbance degree
high
low
high
low
high

2.2 Field census
For the census of the ground-beetles, the pitfall trap
method was set at every 5 plots. It is the method of collecting
the insect by which buries the with 7 cm in diameter, a depthof 10 cm, and a capacity of 280 ml glass made from plastic
materials with no bate on the ground (Fig. 2). 20 traps wereset at every 2 m from each other in each plot, a total of 100 ofa trap buried at a time. The pitfall trap was remained for 2days at one census. The species and the numbers of collecting
the ground-beetles was identified. The recorded beetles were
removed from the traps.
Census was carried out
from August 10, 2010 to
December 7, 2010 atevery week, the total of
23 times census weredone.

3.1 Field census
A total of 3707 ground-beetles, representing 82 sepecies, wascollected in the census in Table 3. Only two species, Colichus
halensis and Harpalus niigataus, were observed in all plots.
Some species were collected only one plots. Especially ingrassland in Shonai biotope 6 indegeous species wereobserved.
The Statistical analysis of the census was shown in Table 4.
The value of diversity index was almost same in all plots. But
the highest Shanon-Winner index was observed in grassland
in Yadagawa biotope. The same result was shown about theevenness of Chao index. From the index of disturbance,
Ishitani index, the forest plots in both biotopes were classified
into forest area, and other plots into the flood plain in Fig. 3.
The result of DCA of the ground-beetles assemblages wasshown in Fig. 4. It indicated that different vegetation plots
placed along the first DCA axis increasingly,
Shonai-grassland < Shonai-wetland < Yadagawa-grassland <Yadagawa-forest < Shonai-forest. It also indicated thatdifferent ecological characteristics placed along the first DCA
axis increasingly, grassland-species < generalist <forest-species.

Fig. 2. The picture of a pitfall trap. .

4. Discussion

ShonaiIt became clear that various the ground-beetles communities
are maintained in two biotopes, Shonai and Yadagawa thathave same river system, by this research. There were greater840

differences in the temporal activity density and species
richness of the ground-beetles among the vegetation types
with disturbance frequency.
Table 3. The list of collected ground-beetles at each plots.

Fig. 3. The relationship between vegetation type and Ishitani
disturbance index. The value of the index in each plot were shownin circles.

Fig. 4. DCA of the ground-beetles. ・: each species of collected in
all plots, circle: the average of each plots.

Table 4. The statistical values in each plots.

4.1 Characteristics of the ground-beetles in Biotopes
In this investigation, many Harpalus sinicusm, Harpalus
niigatanus of phytophagy, habitat generalist Campalitachinense, were collected in the whole region. From this, it canbe said that these kinds are representation kinds of the Shonairiver system in Nagoya. The respectively characteristic
species was collected in each biotope, Pheropsophus
jessoensis, Eusilpha japonica and Damaster blaptoides
blaptoides at Yadagawa biotope, Dolichus halensis andOnthophagus tricornisi at Shonai biotope.
Campalita chinense has flight capability, the capability to
produce a large number of eggs and the resistance to841

disturbance [20]. Dolichus halensis is carnivore with flightability and is widely distributed over whole extent of Japan
from mountain land to urban areas. Harpalus species has
flight capability and eats grasses that distributed widely. This
genus was observed in all the plots as a generalist.
Damaster blaptoides blaptoides is an endemic species of
the area and their food is snails. The distribution of thespecies is widely from a forest to a grassland [21]. Although
Pheropsophus jessoensis has flight capability and an imago iscarnivorousness, the ecology of a larva is specialized, it isparasitic eggs on Gryllotalpidae. The habitat of the species isstable grassland. It is decreasing with reduction in
Gryllotalpidae in recent years. Eusilpha japonica is acommon species that has flight ability and is distributed overmountain land from the flatland. The habitat of the species isa little stable forest. It is thought that the area where these two
species were collected is comparatively stable. The forest
specialist who inhabits more stable mature forest was notcollected. Most forest specialists are indicator species andrare species. As mentioned above, the ground beetles
community of the Shonai river system is consist of the species
that has flight capability, power of resistance to disturbance.
Because of the Shonai biotope was constructed one years ago,the community structure of the Shonai biotope is consist of
disturbance dependence species. On the other hands, it passed
three years after the construction of Yadagawa biotope, thespecies of the community structure is consisted of the species
that likes stable grassland.
Saprinusspl endens is a kind of Histeridae which likes
stable environment[22]. In this research, only one species wascollected in the Shonai river wetland. It did not breed andgenerate in the wetland one year after development, but it isthought that it came flying from other habitats. The sameway, the species of Nepa hoffmanni can be considered themigration from another habitat.
Damaster blaptoides blaptoides, Carabus arrowianus andOnthophaqus tricornis are generalists, that live in the places
where many kinds of vegetation is mixed. The species wascollected all over the plots, then when disturbance happened,
the species evacuates a surrounding grassland area or
woodland area temporarily.
4.2 The cause of the biodiversity in biotopes
Through all the research periods, there were most the
numbers of individuals and the species of the Yadagawa
biotope grassland plot. It was shown that there was a twice asamount of biomass as this by simple calculation. However,there is no difference of diversity index among five plots. It isbecause the collected number of Phropsophus jessoensis
loses the balance of the species composition of each
investigation plot (291 individuals) of the Yadagawa
grassland plot. If the dominant species collected over 40
individuals is removed, the other species are collected with
sufficient balance, became a factor of high biodiversity andshowed the characteristics of each community of theground-beetles.
The result in which the classification result of theinvestigation plots, grassland, forest, wetland, by Ishitani's

disturbance index[17] reflected the feature of theinvestigation plots well was obtained. But only a few
individuals of the species that reflect the wetland, such as
Archipatrobus flavipes, Scarites terricola pacificus, Drypta
japonia, Chlaenius variicornis, and Pterostichus sulcitarsis,
were collected in Shonai wetland plots. And also the speciesof Bembidiinae was not checked. On the other hands, forest
species, such as Haplochiaenius costiger in Yadagawa forest
and Pterostichus noguchii in Shonai forest plot, werecollected. This is considered that the construction time of thehabitat affected the community of each plot.
Some studies indicated the axis 1 of DCA is environmental
fastidiousness [23]-[25], on the other hands the axis 2 of DCA
showed nothing. The same result was shown in this study.
Since the biotopes of this research was recently constructed,
the litters layer has piled up only about 1 cm in depth. Thenthe biotopes are not evacuation habitat yet for the
ground-beetles. It is thought that the biotopes became thecollecting feeds area and laying eggs area. In Shonai bitope
grassland plot the mowing is done completely, but the
production of seeds by Gramineae was observed. There are a
sufficient biomass to maintain the Harpalus community in
Shonai biotope grassland plot. Pheropsophus jessoensis wascollected in grassland plots and wetland plot in both biotopes.
The species has a special feature that larva of the species lives
on Gryllotalpa orientalis. But G. orientalis was not collectedsufficiently in Yadagawa Biotope, then it is not showed thatthe reproduction of the species was occurred. The Adult of thespecies utilize as the feed area, and in the other plots they
breed and laid eggs. In the Shonai biotope collection,
sufficient number of G. orientalis was collected, thereproduction of P. jessoensis was occurred.
It was pointed out that Chlaeniu micans lives in forest andtheir larva feed the caterpillars on the vegetation [26][27]. In
both biotopes C. micans was collected in early autumn. The
reason is that a mowing in September was reduced thecaterpillars, then there was no feeds in the biotopes.
4.3 The conservation of the biodiversity in biotopes
Many researchers have pointed out that the management
method of vegetation has seriously influence on a biologicalcommunity [27]-[31]. Forest specialists can invade the areasthat maintaining the stable forest environment, accordingly
succession degree be established gradually. It can be said thatthe succession degree affects influence on the ground beetles
community also in the riverbed. From the data of Shonai
biotope, completely mowing affects the biodiversity, then it isimportant by leaving without cutting down some willows to
aim at recovery of the ground beetle community which makesthat a habitation base. The immigrant from the habitat of thecircumference of a biotope was suggested. It is important thatstepping stones or corridor to which the import to a biotope isurged to form a habitat network.
5. CONCLUSION
We get two main conclusion.

8421) Investigated the ground beetles community was the sametendency as the structure of the community of a generalriverbed. Although the species that is characterized by thebiotopes was not checked, the ground-beetle community offorest habitat was collected also in small-scale riverbed
forest.
2) In the newly constructed biotopes, the ground beetles
community was reflecting the characteristic of the area, andcan use it for environment assessment as an index whichshows maturity and disturbance degree.
Based on these conclusions, three proposals about the
management method of a biotope are made.
1) In a part of low riverbed area, it refrains from felling,
willows are grown up naturally, and small-scale riverbed
forest is made.
2) Within the distance from which a grassland species canescape, mowing makes a mowing zone and a non-mowing
zone like "tiger cutting", and various vegetations are arranged
like a mosaic.
3) In the tall grassland and riverbed forest like a reed or
willows, the conservation area leave the natural succession is
set up.
6. ACKNOWLEDGMENT
We wish to thank the member of Ministry of Land,
Infrastructure and Transport Shonai River Office for researchsupport and permission to work in the river. We also thank themembers of our laboratory for their assistance in the fieldworks.
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[3] Morishita M, “Measuring of interspecific association andsimilarity between communities” Men.Fac. Sci. Kyushu
Univ. Ser. E. Biol, vol.3, 1959, pp. 65-80.
[4] Inaizumi M, Butterflies in Tochigi Prefecture. Insect
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[5] Tanaka B, “The one method of environment assessmentby butterflies-The recent advances in butterfly
mathematics,” Lepidopterological Society of Japan
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[6] Ishii M, et al., “Diversity of the butterflis community inthe city park in the Osaka Prefecture,” Jap. J. of Environ.
Ent. and Zool. vol. 3, 1991, pp183-195.
[7] Sunose T, “The one method of butterflies community
research,” Ruin and Conservation of butteflies, vol. 2,
1993, pp. 83-90.
[8] Sunose, T, “Environment assessment by line transect
methods” Insect and Nature, vol. 31, 1993, pp. 9-12.
[9] Nakamura H, “Methods of the analysis of the biotic

community and the environmental assessment,” J. of
Facul. of Agri. Shinshu Univ., vol. 36, 2000, pp.1-10.

[10] Ishitani M, “Ecological research on the ground beetles as
an environmental indicator,” Hiwa Sci. Muse. Rep. Res.,
vol. 34, 1996, pp. 1-110.
[11] Thiele HU, Carabid beetles in their environment. An
study of habitat by selection adaptation in physiology
and behavior. Springer-Verlag: 1977
[12] Ishitani M, “ The ecological research of the groundbeetles: different vegetation area affect the community,”
Chugoku area Insect, vol. 17, 2004, pp. 13-18.
[13] Hiramatsu S, “The community of the ground-beetles in
the forest of the Hakusan,” the Research report ofHakusan conservation center, vol. 31, 2004, pp. 55-65.
[14] Niemela J, “Effect of changes in the habitat on carabidassemblages in a wooded meadow on the Aland Islands,”
Notulae Entomologicae. vol. 69, 1990, pp. 169-174[15] Lenski RE, “The impact of forest cutting on the diversityof ground beetles [Coleoptera: Carabidae] in the southern
Appalachians,” Ecol. Entomol., vol. 7, 1982, pp.
385-390.
[16] Maleque MA et al., “Arthropods as bioindicators ofsustainable forest management, with a focus onplantation forests,” Applied Ent. and Zool., vol. 44,
2009, pp. 1-11.
[17] Ishitani M, “The wandering ground beetles as anenvironmental indicator,” Insect and Nature, vol. 31,
1996, pp. 2-7.
[18] Mano T, et al., “Bait trap survey of insects in the central
part of Toyota City (Aichi prefecture),” Res. of Yahagi
river, vol. 10, 2006, pp. 15-42.
[19] Lee C, et al., “Change in structure of ground-beetle
assemblage onleeve of paddy field after the paddy
reclamation in southern Osaka, central Japan,” Insect andNature, vol. 43, 2008, pp. 6-10.
[20] Ishitani, M Yano K, “Species Composition and seasonal
activities of ground beetles (Coleoptera) in a fig orchard.
Jpn. J. Ent., vol. 62, 1994, pp. 201-210.
[21] Ishitani M, “The biology of the ground beetle - things thestart-,” Insect and Nature, vol. 43, 2008, pp. 2-5.
[22] Mochizuki S, ”Studies on ecology of the bright blue
hister beetle, Saprinu speciosus Erichson (Coleoptera:
Histeridae),” Jpn. J. Appl. Ent. and Zool., vol. 29, 1985,
pp.21-25[23] Tanaka K, “The water assessing-environment method bybiota,“ The agricultural environmental monitoring
manual for water environmental preservation, second ed.
National Institute for Agro-environmental Sciences:
2005, pp. 211-216..
[24] Kagawa Y, “Studies on structure and conservation

biocontrol use of ground beetle assemblages inagricultural landscape” in Kobe University repository
thesis, 2005.

[25] Seyfferth A, “Diversity of ground beetles and saproxylic
beetles (Coleoptera: Carabidae Div. Saproxylic) in East
Mediterranean ecosystems (Israel): dispersal, habitat,
activity and reproduction,” ine-book,
urn:nbn:de:gbv:luen4-opus-141889 URL:
http://opus.uni-lueneburg.de/opus/volltexte/2010/14188/[26] Tanaka B, et al., “The insect fauna of between Heisei-M
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Yahagigawa River,” Res. of Yahagi river, vol. 2, 1998,
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[27] Matsumoto K, “Management of the village-vicinity
mountain, and the biodiversity,” Insect and Nature, vol.
43, 2008, pp. 20-26.
[28] Suzaki T, Tanaka B, “Examination of mowing method
which supports the inhabits of plants, insects and spiders
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[29] Tanaka B, “Insects in the researches along the riversideof the Yahagigawa in urban blocks of Toyota City from1995 through 1999,” Res. of Yahagi river, vol. 5, 2001,
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844

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Dense granular-fluid mixture: the effects of the natural grains characteristics
and the interstitial fluid viscosity
Anna Maria Pellegrino and Leonardo Schippa
University of Ferrara, Italy

ABSTRACTAn experimental study about the rheology of debris flow was carried out on debris flow soils and on natural sand using
several rheometrical tools in order to investigate the rheological behaviour of the “fluid-like” material mixture. The
results obtained on debris flow material mixtures suggest that, in the fluid-like regime, the typical rheological behaviourof this mixture is that of yield stress fluids and the rheology of the mixtures is strongly dependent on the grain
concentration. The velocity profile obtained on natural sand-Newtonian fluid mixtures were analyzed in order to
identify the shearing zone and to explain the dependence of the flow characteristics (i.e., transition from quasi-static
regime to liquid regime) on viscosity and shear rate. The results suggest that, in that field, the interstitial fluid viscosity
influence the sharing material layer and that no flow is possible for solid fraction higher than a maximum value.
Keywords: Dense granular suspensions, rheology, solid concentration, interstitial fluid viscosity, velocity profile

1. INTRODUCTION
In the last 50 years, the risk of landslides has increased
enormously. The soil characteristics have already been
explored, but there is still a lack of studies focused on theproperties of the ‘post-failure’ material. Within the frame of a
fluid mechanics treatment, the rheological characteristics ofthe flowing material would be useful for a better
understanding of the flow characteristics of such events. In
this paper, interesting informations in that field have been got
by studying the rheological characteristics of a debris flow
dry soil remixed with different amounts of water and a natural
monogranular sand remixed with Newtonian fluids having
different viscosity. These are materials that may behave like ayield stress fluid and their rheological parameters are strictly
related to grain characteristics (i.e., size, shape, etc) and tointerstitial fluid viscosity [1]-[2]-[3]-[4]-[5]-[10]-[11]-[13].
Because of the wide grain size distribution of the debris flow
material (varying from clay to gravel), it has been prefer torefer as the solid volumetric concentration, which is aparameter that relates the solid amount to the total volume ofthe entire mixture. On one hand, with a conventionalrheometer it is possible to analyze the rheological behaviourof the debris flow material considering a 70% of the wholegrain size distribution and to study the changing of therheological parameter with the variation of the solid fraction.
On the other hand, with a large scale annular rheometer it ispossible to analyze the rheological behaviour of a large
particles mixtures (i.e., the monogranular-sand suspensions)and the influence of the grain size and of the interstitial fluid
viscosity on different characteristic flowing parameters.

2. EXPERIMENTAL SETUP, PROCEDURES AND
MATERIALS TESTED
A rotational rheometer CVOR (Bohlin Instruments) equippedwith a vane rotor geometry system had been used, which
consists of four thin blades arranged at equal angles around asmall cylindrical shaft. The blade radius was R1=13 mm, andthe blade height was L=48 mm. The vane rotor was immersedin the sample (the sample volume was roughly 27 ml)
contained in a cylindrical cup of radius R2 = 18.5 mm. During
the test, a part of the material is trapped in the blades so that asa first approximation the flow characteristics are similar to
those between two solid coaxial cylinders, with the inner
radius equal to that of the blade [12]. Under usualassumptions (no inertia effects, negligible normal stressdifferences) one can estimate the shear stress τ and the shear
within the material from the following formulae:rate

T / 2 R12L

R1 / R2

(1)

R1

(2)

It was possible to define the deformation undergone by thematerial from the initial time. At time t, the deformation γ isequal to:
t

dt

(3)

0

Creep tests were carried out, which consisted of imposing a
constant stress and measuring the induced deformation of thematerial over time. The deformation is expressed in terms of845

angle of rotation of the inner cylinder since the initial
application of the stress. A complete test consists of imposing
various values of stress, preparing each sample in the same
initial state before each new stress value.
A large scale annular rheometer has also been used, which
consists of two concentric horizontal aluminium circulardisks assemblies mounted on a vertical shaft fixed on thebasement of the apparatus. The lower disk assembly ismounted on the shaft with two roller bearings allowing
rotation but no vertical motion. The lower disk is restrained
from rotating by a torque arm connected to a load cell. This
lower part has an annular trough 100 mm wide and 35 mm
deep and has a mean radius of 200 mm; it is therefore called
channel. The channel is closed above by a heavy rotating
upper disk assembly mounted on the same vertical shaft
through two roller bearings allowing both rotation andvertical movements. The surfaces of the upper (i.e. the ceilingof the channel) and of the lower disk (i.e. the bottom of thechannel) were made rough with a glued sand paper sheet
having a roughness of about 1.2 mm. The angular velocity is
obtained by recording the pulse rate and, knowing the size ofthe gear and of the cogged wheel, the rotation speed of theupper plate of the shear cell is easily obtained. The verticaldisplacement of the upper disk is measured, with respect tothe channel bottom, via a cable linear position transducer.
Before starting a run, the shearing channel is filled with aknown amount of granular material, carefully mixed with theliquid (special attention is paid to removing any trapped air).
After a few rotations aiming to settle the material (the lower
plate remaining at rest), the vertical position of the disk ismeasured. The counter weight strictly necessary to balance
the disk weight is determined and, by unloading thecounterweight, a fixed vertical pressure is caused. The disk isthen let turn and a run is performed according to one of thetwo following procedures: one or more assigned velocities of
the upper disk are reached, and the corresponding upper disk
displacements are measured; the velocity is increased until adesired displacement of the rotating disk is reached; thevelocity of the upper disk is increased step by step andsubsequently decreased as well. During the test several
desired velocities are reached. Torque, velocity anddisplacement are anyway monitored and a video record istaken.
The debris flow material has been tested with the rotationalrheometer. It was sampled from the source area of a debris
flows occurred in the Campania region (Nocera, Salerno,southern Italy, March 2005). The soil type, in a thickness ofabout a meter, depends on the most recent pyroclastic
deposits deriving from the volcanic activity of Mount
Somma/Vesuvius. The grain size distributions of thecollected sample are reported in Fig. 2.1: the soil is a sandy
silt with a small clay fraction. The bedrock underlying the soil
is limestone, the specific gravity of soil particles is 2.61, thedry and total weight of soil per unit volume are respectively
9.08 and 11.35 kN/m3 and the porosity is 0.66. With thedebris flow material, it was decided to perform tests on thesoil fraction with a particle diameter less than 0.5 mm, inorder to respect at best the continuum assumption (it isnecessary to have a material thickness much larger than theparticle size in the material). Doing so retains about 50–70%
of the whole grain size distribution, as shown in Fig. 2.1.

Fig. 2.1 Grain size distribution of the collected debris flow
material.
All the experiments involving debris flow materials werecarried out with mixtures of dry soils with different amountsof water. The solid concentration is described with the help of
the solid volume fraction Φ, that is, the ratio of the volume ofsolids, VS, to the total volume VT (water plus solids) of thesample. In fact, according to the mixtures involving debris
flow material and water, it is not possible to determine exactly
how much of the dry material contributes to the formation ofinterstitial fluid (i.e., composed by water and smaller
particles) and how much of the material set up the solid phaseof the mixture (see Fig. 2.2). For each material tested,material mixtures of about 500 ml were prepared, mixing
soils and distilled water with an electronic mixer (at about 30
rev/min) for 15 min. Then a sample volume of about 30 ml
was used for each test at a constant temperature (23°C).

Fig. 2.2 Tested materials. a) debris flow material; b)natural sand.
The natural sand has been tested with the annular rheometer.
The natural sand have uniform granulometric distribution(d=1.4-1.68 mm), the shape of the grains is irregular and thedensity is equal to 2670 Kg/m3. All experiments were carriedout with mixtures of dry sand remixed with different fluids:
clear water (low viscosity interstitial fluid, 0,001 Pa·s) and.
Glycerine (high viscosity interstitial Newtonian fluid, 0,394
and 0,598 Pa·s). The combinations of fluid and grains lead to
mixtures of fluid and irregular grains having solid-fluid
density ratios ranging from 1 to 3. According to [13] thelinear concentration λ, i.e. the ratio between particle diameterand their relative distance, related to the actual concentration
Φ and to the maximum theoretical bulk concentration Φ0 (i.e.
the concentration at which all particles are in reciprocal
contact) has been used. As reference concentration, the846

maximum concentration Φm = 0.56 has been used. The actual
concentration has been derived from the displacement of theupper disk. The residual volume was measured and taking
into account the residual volume and the corresponding
displacement of the upper disk, the granular concentration of
the shearing layer is then derived. It is important to note thatin the case of the natural sand has been possible to accurately
determine the maximum packing concentration (for thedetailed procedure see [4]) while for the debris flow material
it has been simply deduced.
3 EXPERIMENTAL RESULTS
3.1 Creep tests
Creep tests were carried out imposing a constant stress to thedebris flow material mixtures and measuring the induced
deformation over time. A complete test consists of imposing
various values of stress, lower and larger than a critical stress,
called static yield stress (i.e. the value of stress after that the
mixture starts to flow like a fluid) [9]. The value of the static
yield stress for each debris flow material mixtures analyzed
was determined with some stress sweep tests carried out using
the rotational rheometer. The stress sweep tests consist inmeasuring the apparent flow curves by applying an increasing
shear stress ramp. The shear stress was continuously
increased in a logarithmic way from 0.1 Pa to a large(maximum) value, and the corresponding shear rate wasmeasured. The maximum stress value was specificallydetermined for each sample from independent (rheometrical)
tests under similar technical conditions so as to get a rapid
flow in the liquid regime without fluid expulsion out of thegeometry (for the complete procedures and results see [11]).
Before each creep test performed, the material was set up
inside the geometry, immediately imposing a pre-shear at the
maximum stress value above described for 30 s, and thenstarting the test in order to homogenize the sample. In Fig. 3.1
is showed the experimental results obtained with a creep test
on debris flow material mixture at solid volumetricconcentration Φ of 42%. It is observed that for stresses lower
than critical stress the deformation’s curve remains concave,with a slope continuously decreasing in time on a logarithmic
scale, and exhibits an apparent horizontal asymptote. The
deformation is limited, and the instantaneous shear rate
decreases continuously to lower values, so that no steady stateis reached. Finally, the material apparently stops moving andthis should correspond to the solid regime of the material (theapparent shear rate
is close to zero) [14]. For stress values
higher than the critical value, the initial slope of the curve issimilar to that under smaller stresses, but after some time
there is an inflection point, and the curves gradually tend toreach an inclined straight line, with a slope equal to 1. In thiscase the deformation increases at a constant rate, whichmeans that the apparent shear rate is constant, and asteady-state flow has been reached. These curves correspondto the liquid regime of the material [15]. However, thetransition from the solid to the liquid regime is abrupt (seeFig. 3.1): for a stress just above the yield stress, the material
rapidly reaches a shear rate level that is relatively high(typically of the order of or larger than 1 s -1). This suggests
some viscosity bifurcation effect [9], again associated with acritical shear rate ( c ) below which no steady flow can beobserved, as has already been observed from the results
obtained in the sweep test on the same material mixtures.
Note that the apparent critical shear rate may depend on theflow history, which can explain a slight difference between
the value deduced from Fig. 3.1 and that found from thespecific procedure used for sweep tests, showed in Fig.3.2. A
recent work [8], focused on the behaviour of concentrated
granular suspensions exhibiting similar characteristics,
showed that this material exhibits a critical shear rate below
which they cannot flow steadily. This could be an effectleading to shear-banding when an apparent shear rate is
imposed below this value. The same trends occur here withthe debris flow material mixtures analyzed: the static yieldstress corresponds to the stress needed to unjam the locallysettled structure, and then the dispersed suspension flows
more easily. Consistent with this scheme, along the stressplateau associated with the static yield stress the material
shear-bands and, increasing shear rate beyond the criticalvalue, the material flows as a homogeneous fluid. (seeFig.3.2). As a consequence we have an unstable behaviour:when a stress is imposed while the mixtures are at rest they
will not flow until the static yield stress has been reached, andjust beyond this value suddenly flow at a shear rate larger thanthe critical shear rate. It has been noted that, in the absence of
sedimentation and with very small particles, the material
behaves like a Newtonian fluid up to a critical shear rate[6]-[7]-[8] - [9].

Fig. 3.1 Creep tests for debris flow material (42%).
Deformation against time for different stress levels (frombottom to top): 90, 110, 120, 140, 142, 160, 170, 180, 200
Pa.

Fig. 3.2 Sweep tests with debris flow material 1 at 42%(Representative test).

847

new set of measurements is carried out along a different
vertical. Measures along different verticals, located at
different positions apart from the wall, were repeated as manytimes as possible depending on the optical characteristics ofthe mixture. The most remote measuring point was usually
located 3-4 mm apart from the sidewall towards the core of
the channel (about 2-3 grain diameter).

Fig. 3.3 Rheological characteristics of debris flow
material as a function of solid fraction: static yield stress(empty squares symbols) and critical shear rate (empty
circles symbols).
This behaviour is strictly related to the characteristics of thesemixtures, i.e. particle type and interstitial fluid viscosity,
which are strongly related to the solid volumetricconcentration of the mixtures as showed in section 2. In fact it
is especially interesting to look at the variations of the static
yield stresses and the critical shear rate with the solid fraction(see Fig. 3.3). It can be seen that the yield stress and thecritical shear rate increase in a similar way with solid fraction
and they increase by two orders of magnitude over the very
narrow range of concentration analyzed(less than 10%) in
which the debris flow material mixtures tested behave ashomogeneous fluids (for more details on the identification of
the range of concentrations in which this type of material may
be considered as fluids and thus can be characterized with theusual rheological tools see [11]). It is important to note that if
the critical shear rate increases more or less like the yieldstress when the solid fraction increases, then the material
strength increases, and its apparent velocity when it starts toflow increases proportionally.
3.2 Velocity profiles
Measurements of velocity profiles were performed by Laser
Doppler Anemometer equipment mounted on the large scaleannular rheometer [4]. The used system is made up of a Flow
Lite integrated laser-optics unit with a Flow Velocity
Analyzer signal processor (FVA). The optic unit is mounted
on a specific support, and it is driven manually in order toperform measurements at different point inside the shearing
channel. Typical run was performed in analogy with theformer procedure, i.e.: the channel is filled by a known
amount of grains and fluid, by a counterweight a normal load
is applied to the rotating upper disk, after few rotations theupper disk remains at rest aiming to settle the material andextrude any trapped air,. the vertical displacement of theupper disk is measured, the upper disk is set into rotation at
the fixed velocities and measurements are then accomplished.
In every run, the entire channel depth is visible through thewindow mounted on the external wall. Velocity
measurements are performed inside the shearing gap along a
vertical at different levels. The first set of measurements deals
with a vertical just close to the wall. Afterwards, themeasuring volume of LDA is moved inside the channel and a

848

Fig. 3.4 Torque vs. the rotational velocity in the steadystate of a 56% suspension. The line is a fit to a Bingham
model: T=Tc + αΩ with Tc=3.86 N m and α=0.53 Pa·s.
In this way any boundary effect related to the presence of thewall (which was actually evident), could have been evaluated;both velocity profiles and actual shearing zone could havebeen extrapolated, as it is described in the following.
Measurements were carried out during several runs,
reproducing with strict approximation the experimental
condition of the corresponding rheometer tests; velocity
profiles were measured along few verticals from the external
wall towards the centre of the channel. In Fig. 3.4, the torquevs. the rotational velocity for the steady flows of a 56%
sand-water suspension has been reported. As observed in theexperiments on debris flow materials (see previous section), it
has been observed a quasi static regime (i.e., the solid regimeof the material) followed by a viscous regime (i.e., the liquid
regime of the material). It has been found a shear torqueplateau at low velocities and a linear increase of torque withthe velocity above a critical velocity Ωc (Fig. 3.4). The data
can be well fitted to a Bingham law T=Tc + αΩ with Tc=3.86
N m and α=0.53 Pa·s: this suspension seems to exhibit a yieldstress and to behave macroscopically like yield stress fluids.
In the following, the regime for which the torque increases
linearly with the rotational velocity will be called the“macro-viscous” regime (as in [13]). A typical measured
velocity profiles obtained are plotted in Fig 3.5: thedimensionless velocity profiles for the steady flows of a 56%
sand-water suspension for various rotational velocities. As it
is evident the measured velocity profiles are similar in shape.
Measurements show that the velocity profiles are roughly
exponential and that they occupy the whole gap for theconsidered rotation velocities. These velocity profiles show
the mixture in the called macro-viscous regime [4]-[13], inwhich typically the applied shear rate is higher than thecritical one. In fact, in this regime (i.e. at high rotation

velocities), the mixture inside the gap is all sheared. In thecase, the whole sample is sheared, and all the reduced
velocity V(zi)/V(z) profiles plotted vs. the radius fall along
the same curve. It is noted that the critical velocity below
which the shear flow is localized is as the same order ofmagnitude of the one associated with the torque plateau in
Fig. 3.4. The velocity profiles obtained confirmed the Non
Newtonian character of these mixtures: indeed the velocity
profiles in the macro-viscous regime are very different fromthose expected for a Newtonian fluid; i.e. a quasi-linear
velocity profile. In Fig. 3.6, it has been plotted thedimensionless velocity profiles of sand in a 56% suspension
with different interstitial fluids at equal rotational velocity (20rpm). This rotational velocity belongs to the macro-viscous
regime of these three suspensions, in which the mixtures
analyzed behaviour like a fluid and easily flow. In the case of
the water-sand suspension, it has seen that in themacro-viscous regime all the gap is sheared. In the case ofglycerine-sand suspensions, it has been observed that the gap
cannot be fully sheared: there still is a 1 mm region where thematerial is not sheared. In this case, when the rotational
velocity increases, the thickness of the sheared layer does notincrease anymore for Ωi>Ωc, and it remains smaller than the
gap size. Analyzing velocity profiles and video records, it
seems reasonable to assume an active shearing layer zm,
which extends from the ceiling of the channel to the level
corresponding to 10% of the higher velocity (see Fig.3.7).

Fig. 3.5 Dimensionless velocity profiles in the steady stateof a 56% water-sand suspension, at various rotational
velocities.
Below this level we assumed that the quasi-locked lower
layer zs is at rest .It has been thus observed that the thicknessof the sheared layer in the macro-viscous regime (i.e. forΩi>Ωc), does not depend on the velocity and cannot exceed amaximum value zm, which decreases when the suspension’s
viscosity increases. A value of 7.5 mm was found for thewater-sand suspension, whereas 7 mm for the glycerine-sandsuspensions. Actually, we observe that viscous ﬂows at low
velocities are not stable (i.e. for Ωi>Ωc andC ), and thatthe ﬂow localizes. The mechanical origin of localization wasevidenced by [7]: there is no steady ﬂow for a shear rate
below a critical shear rate C that depends on the material’s

849

properties. When the rotational velocity is lower than Ω c, theapparent shear rate in the gap is lower than C . Therefore,
the ﬂow has to localize so as to ensure that in the ﬂowing
material
C and during localization, the shear stressincreases whiletends toC : the stationary shear stress at

the walls is then approximately
thus leads to a shear stress plateau

C . Shear localization
C

at low velocities (Fig.

3.4); however the ﬂowing behaviour of the sand mixtures
analyzed is always a purely viscous behaviour, even when theﬂow is localized, in the range of velocities studied. Finally, it
has been noted that the mixtures behaviour dependences on
the volume fraction: there may be no ﬂow for Φ m. The
macro-viscous regime of dense suspensions then starts when
the sheared region has reached a region where Φ=Φ m: thethickness of the sheared layer cannot increase anymore, andnow the torque has to increase linearly with the rotational
velocity since the local shear rate now increases linearly withthe velocity and since the ﬂowing material behave like apurely viscous material. It is finally showed in Fig. 3.8 thecomparison between the variations of the yield stresses withthe solid fraction for every material mixtures tested in thisstudy. It can be seen the same trend regardless the material:
yield stress increase with solid fraction in exponential way
and debris flow material mixtures and water sand suspensions
increase in a similar way with solid fraction.

Fig. 3.6 Dimensionless velocity profiles in the steady stateof sand in a 56% suspension with different interstitial
fluids at equal rotational velocity (20 rpm).

Fig. 3.7 Sketch of shearing and quasi-locked layer in theshearing gap.

5 REFERENCES

Fig. 3.8 Static yield stress as a function of solid fraction
for all the material mixtures tested: comparison between
debris flow material and monogranular sand.
4 CONCLUSION
In this paper it has been studied the flowing behaviour ofdebris flow material-water mixtures and natural sand waterand glycerine suspensions by coupling local velocity andconcentration measurements through large scale annularrheometer and macroscopic rheometrical experiments in avane geometry. From rheometrical experiments performed
and with the help of the velocity proﬁle measurements, it has
thus shown that the local material behaviour is a purely
viscous behaviour. Creep tests and sweep tests have beenshown that the typical rheological behaviour of thesesuspensions is that of a yield stress fluid exhibiting a static
yield stress. The flow of such materials is usually unstable:
they will start to flow beyond a critical stress, but just beyond
this value will reach a high shear rate associated with a highflowing velocity. Then, the yield stresses and the critical
shear rates are strongly influenced from solid fraction: if thesolid fraction increases they increase in an exponential way.
Analyzing velocity profile has been possible to identify andquantify the actual shearing layer and deriving the relatedparameters. Velocity measurements have been confirmed thatthese kind of mixtures are non Newtonian (they can flow only
if a critical value of velocity has been reached, i.e., critical
shear rate) and that the grain characteristics not appear toaffect the macroscopic rheological behaviour of both
mixtures. Moreover in the velocity profiles it has been clearlyseen the influence of the interstitial fluid viscosity on the
mixtures behaviour: the shared material layer decrease whenthe viscosity of the interstitial fluid increase and the relative
velocity also decrease. These results suggest that, in the field,a small change in solid fraction will cause a slight decrease of
the static yield stress, inducing a flow rapidly reaching a shear
rate larger than the critical shear rate associated with a rapid
flow. These results raise many questions: first, the physicalmechanism associated with critical shear rate; second, thecause leading to the formation of two distinct layers within
the material: the lower quasi-locked layer and the upper fully
sheared layer; third, the characterization of the suspension atvery low velocities in the quasi-static regime.

850[1] Phillips CJ & Davies TRH. Determining rheologicalparameters of debris flow material. Geomorphology 4,
No. 2, 1991, pp. 101–110.
[2] Major JJ & Pierson TC. Debris flow rheology:
experimental analysis of fine-grained slurries. Water
Resour. Res. 28, No. 3, 1992, pp 841–857.
[3] Coussot P & Piau JM. A large-scale field coaxial
cylinder rheometer for the study of the rheology ofnatural coarse suspensions. J. Rheol. 39, No. 1, 1995, pp.
105–124.
[4] Schippa L & Lamberti A. Experiments on the rheology
of a dense granular flow. Final reposrt of EC Research
Programme DFR ENV4-CT96-0253. June 1999, pp.
1-34.
[5] Schatzmann et al.. Rheometry for large-particulated
fluids: analysis of the ball measuring system andcomparison to debris flow rheometry. Rheol. Acta 48,
No. 7, 2009, pp. 715–733.
[6] Ancey C & Coussot P. Frictional-viscous transition forconcentrated suspensions. C.R. Acad. Sci., Paris 327,
1999, pp. 515–522.
[7] Huang et al.. Flow of wet granular materials. Phys. Rev.
Lett. 94, No. 2, 2005.
[8] Ovarlez G, Bertrand F & Rodts S. Local determination of
the constitutive law of a dense suspension of
noncolloidal particles through MRI. J. Rheol. 50, No. 3,
2006, pp. 259–292.
[9] Fall et al.. Yield stress and shear-banding in granularsuspensions. Phys. Rev. Lett. 103,No. 17, 2009.
[10] Pellegrino et al.. Rheological behaviour of pyroclastic
debris flow, 3rd Int. Conf. “Debris Flow III”, Milano,
Italia, 24 - 26 May, 2010.
[11] • Scotto di Santolo et al. Rheological behaviour ofreconstituted pyroclastic debris flow, Geotechnique
Journal, Vol 62, Issue 1, 2011, pp. 19 –27.
[12] Nguyen QD & Boger DV. Direct yield stressmeasurement with the vane method. J. Rheol. 29, No. 3,
1985, pp. 335–347.
[13] Bagnold RA. Experiment on a gravity-free dispersion oflarge solid sphere in a Newtonian fluid under shear. Proc.
R. Soc. London Ser. A 225, No. 1160, 1954, pp. 49–63.
[14] Coussot P. Rheometry of pastes, suspensions andgranular materials: Application in industry andenvironment. New York: John Wiley & Sons, 2005.
[15] Coussot P. et al.. Aging and solid or liquid behavior inpastes. J. Rheol. 50, No. 6, 2006, pp. 975–994.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Back-analysis of Frictional Jacking Forces Based on
Shear Box Testing of Excavated Spoils
Choo Chung Siung1 and Dominic Ong E. L.2
Swinburne University of Technology (Sarawak Campus), Malaysia

1,2

ABSTRACTTrenchless technology has become the preferred technique for installation of buried infrastructure due to advantages
such as minimised ground surface disturbances as well as reduced social disruptions. These benefits have led to anincreased demand for trenchless technologies such as the pipe-jacking method, with its application well-suited forurban environments. The construction of trunk sewers in Kuching, Sarawak has been carried out using the pipe-jacking
method. Critical to this micro-tunnelling technique were the associated frictional forces anticipated during the jacking
process. Excessive jacking forces could damage installed pipes. The Authors have studied the use of a few
well-established empirical equations to validate frictional resistance during jacking supplemented by laboratory direct
shear box tests carried out on spoils obtained from pipe-jacking drives. The tested spoils were by-products of the slurrytransport system for spoil removal at the tunnel face. Shear box test results have been attempted for use to predict thepipe-jacking forces based on the well-established empirical equations. The predicted jacking forces for this study show
fair agreement when benchmarked against actual, measured forces. However, it is the intention of the Authors to carryout more tests to further verify the reliability of this method for a wider variety of local rock types, especially ingeologically ‘young’ weathered rock where competent rock cores are usually difficult to obtain.
Keywords: Pipe-jacking, direct shear, back-analysis
1. INTRODUCTION
There has been a growing need for trenchless technology forthe construction of various pipelines. This trend has been
evidenced by its implementation in the installation of trunk
sewers in Kuching, Sarawak. Sewer lines comprising of
concrete pipes were installed using micro-tunnelling by
pipe-jacking. This method was elected over conventional
open trench methods due to the minimised socio-economic
disruption and pollution. However, the technicalities of suchtunnelling works required specialist expertise. During thejacking process, various components were necessary for theworks to proceed. The capacity of the microtunnel boringmachine (MTBM), the application of lubrication, the capacityof main jacks, the use of intermediate jacking stations, and thedesign of jacking pipes are constituents of the synchronous
system for successful pipe-jacking. A key factor affecting theabove components is the frictional jacking load.
2. BACKGROUND
Studies have been carried out to investigate the behaviour ofjacking loads during the pipe-jacking process. Pioneeringstudies on pipe-jacking were carried out by researchers at
Oxford University, U.K. Norris & Milligan [1] detailed drives
entailing instrumented concrete jacked pipes were carried out.
Instrumented pipes were outfitted with pore pressure probes,
contact stress transducers, extensometers, and othermonitoring equipment. Data on variations of normal stresses
and pore pressures on the instrumented pipes provided
valuable early insight into the effects of lubrication on jacking

stresses [2, 3]. Pellet-Beaucour & Kastner [4] studied theinfluence of various parameters on the build-up of frictional
forces for drives in varying soil conditions across France. The
authors proposed two frictional force models in their analyses.
Staheli [5] studied the behaviour of pipe-soil interface friction
by conducting laboratory shear box tests. Sand specimens
were applied normally on pipe crown segments of varying
roughness. A similar test was carried out by Shou et al. [6],with the addition of lubrication. Interface frictionalcoefficients were derived from the shear box tests andverified using three-dimensional finite element modelling.
The experiences and results reported by these researchers
have been carefully reviewed and understood, so as to ensurethat the methodologies reported in this paper are within
acceptable conditions and practice.
3. SHEAR BOX TESTING OF EXCAVATED SPOILS
2.1 Direct shear test and geology
Due to the weathered nature of the ‘young’ sedimentary andmetamorphic rock of the Tuang Formation in the Kuching
City area, intact and competent rock cores (≥100mm inlength) are usually difficult to obtain during Soil
Investigation works, thus making the strength parameters (c’and φ’) for these weathered rocks rather elusive. Rock Quality
Designation (RQD) may not be the best indication of rockstrength as the values are usually zero or close to zero due toits friableness. However, this observation does notnecessarily represent the in-situ strength of the rock mass as awhole. Hence, performing direct shear tests where excavation
spoils are reconstituted with appropriate confining stress851

during testing may be a worthwhile and feasible method.
Direct shear test was conducted using the GeoComp
ShearTrac-II direct shear system (Figure 1). The ShearTrac II
system enables full automation of shear box testing.
Automation of the test ensures repeatability of testing regime.

2.2 Collection of excavated spoils
Excavated spoils were collected from the slurry separation
plant (also known as desander). The excavated spoils weretransported from the MTBM face through suspension in apressurized slurry transport system and discharged in thedesander. The desander recycles the slurry through filtrationand segregates the spoils from the slurry. The slurry is thenpumped back through the slurry supply line, back to the TBM
face. Figure 2 shows approximate chainages at which
excavated spoils were collected for testing.
2.3 Shearbox testing and results

Figure 1 ShearTrac II direct shear system
Control of the ShearTrac II system is achieved through theuse of the SHEAR software. Users are given control over the
shear box testing regime by inputting the desiredconsolidation stresses and shear strain rates. Results fromtests are generated regularly in terms of horizontal andvertical displacements, as well as shear stresses along theshear band. The ability of the user to obtain results at ahigh-resolution allows for close scrutiny of tests. The
ShearTrac II system is also able to maintain the consolidationand confining pressures throughout the test, from theconsolidation stage through to the shearing stage.

Test samples were prepared according to AS1289.6.2.2-1998[7]. Scalping of samples was carried out according to theallowances stated in the above-mentioned standard. The test
samples consisted of grains passing the 2.36mm sieve, andretained by the 75μm sieve. Such scalping method has alreadybeen carried out by Bagherzadeh-Khalkhali & Mirghasemi[8]. Scalping was deemed necessary as preliminary shear box
testing showed large fluctuations, particularly beyond thepeak stress threshold. The consolidated drained shear box
tests were conducted with samples placed in the reservoir
filled with distilled water. Shearing was initially carried out atthree confining pressures σ 1 , i.e. 170kPa, 180kPa, and
190kPa. These were approximate in-situ stresses acting on thepipe. Two further sets of tests were conducted at confiningpressures of 100kPa and 250kPa, for better data distribution
when measuring the respective strength characteristics.
Table 1 Results from shearbox testing on excavated spoils
SamplenumberSample 1

Tunnel progress points at
which spoils were collected

Sample 1

Metagraywacke

Receiving shaft

9.5m

Jacking shaft

Silty clay

Sample 2

Sample 2

38m
78m

Sandstone

103m*Not to scale
Figure 2 Estimated locations where Sample 1(metagraywacke) and Sample 2 (sandstone) werecollected as excavation spoils from slurry

σ1
(kPa)
100
170
180
190
250
100
170
180
190
250

Peak strengthcharacteristics
τ peakϕ peakc peak
(kPa)
(°)
(kPa)
117
179
171
36
49
186
227
183
282
282
49
78
282
360

Residual strengthcharacteristics
τ res
ϕ res
c res
(kPa)
(°)
(kPa)
96
128
128
28
37
134
178
112
201
203
47
4
204
275

Table 1 above tabulates the results from shear box tests on the
spoils. Generally, higher confining pressures, σ 1 resulted inlarger shear strengths, τ peak and τ res . Strength characteristics
were interpreted according to the Mohr-Coulomb failure
criterion. Sample 2 showed higher strength characteristics
than Sample 1 at both peak and residual strains. This is
attributed to the particle shapes of the respective samples,with Sample 2 exhibiting generally angular particles ascompared to the rounded particles of Sample 1 (Figure 3). As
expected, peak strength parameters were generally higher
than the residual counterparts for both samples. This isparticularly evident for Sample 1. For Sample 2, the decreasein strength from peak to residual states was significantly due
to a loss of cohesion. There was minimal observed decrease infriction angle of Sample 2. The residual strength of the852

samples is thought to mimic the excavation process better as
the in-situ rock mass is expected to be crushed till failure by
the roller cutter bits found at the face of the MTBM. Besides,
the continuous jacking process involved in the installation ofpipes is also analogous of a shearing process in large–strain
condition. The test results obtained were subsequently
utilized for the prediction of frictional jacking forces.

Table 2 Summary of various parameters for use withfrictional model by Pellet-Beaucour & Kastner [4]
b

δ

K


 π φ De 1 + 2 tan −  
 4 2 ϕ

1

Table 3 Friction coefficient values for various friction
states from Stein (as summarized in Ref. [4])Concrete on gravelor sandConcrete on clay
Asbestos orcentrifugedconcrete on gravelor sandAsbestos orcentrifugedconcrete on clay

Figure 3 (left) Sample 1 (spoils of metagraywacke); (right)
Sample 2 (spoils ofsandstone) used for direct shear tests
3. ANALYSIS OF FRICTIONAL JACKING FORCES
In the pipe-jacking process, jacking forces comprise of soilpressures acting on the face of the MTBM and normal
stresses accumulating along the outer peripheral surface ofthe pipeline. The frictional component nominally forms the
major portion of the jacking loads as the pipe-soil contact area
increases with jacking progress.
3.1 Frictional jacking force equation
Frictional forces have been the subject of much research. The
work carried out by Pellet-Beaucour & Kastner [4] employed
the model shown in Eq.(1), which is expressed as a functionof vertical soil stress on the pipe crown, σ EV as in Eq. (2).

γDe 
γD 

 + K  σ EV + e  (1) σ EV +
2 
2 
2 

h
− 2 K tan δ b(γ − 2bC ) 1 − e b(2)
σ EV =

2 K tan δ F = µLDeπ 

L is the length of the pipe string, D e is the outer diameter ofpipe, γ is the soil unit weight of ground cover, h is the heightof cover at the pipe crown, K is the coefficient of lateral soilpressure, δ is the angle of wall friction in the shear plane, b is
the influencing width of soil above the pipe. μ is thecoefficient of soil-pipe friction, expressed as(3)µ = tan δPellet-Beaucour & Kastner [4] suggested upper limit values δ= ϕ, and lower limit values of δ = ϕ⁄2 or ϕ⁄3 for soil-structureinteraction calculations. They also reported values for b, δand K by various organizations (Table 2).
Staheli [5] and Pellet-Beaucour & Kastner [4] further
reported the use of some values of friction coefficient, μ.
These values were obtained from Stein et al. [9].

Staticfriction

Slidingfriction

0.5 – 0.6

0.3 – 0.4

0.3 – 0.4

0.2 – 0.3

0.3 – 0.4

0.2 – 0.3

0.2 – 0.3

0.1 – 0.2

Fluid friction(lubrication)

0.1 – 0.3

3.2 Case study for back-analysis
A case study was considered for a drive negotiating throughthe geology of the Kuching inner city area. The pipeline lies
embedded in rock, with the crown at a level 9.5m below
ground surface. The 103m alignment of 1.2m diameter
concrete pipes traversed metagraywacke (metamorphic) andsandstone synonymous with the Tuang formation. As
reported by Tan [10], the metagraywacke encountered in the
Tuang formation is characterised as weathered, withoccurrences of quartz veins. This was consistent with findings
from soil investigation works. Metagraywacke in the Tuang
formation is often found interbedded with moderately
weathered phyllite.
The tunnel was excavated using an Iseki Unclemole SuperTCS 1200. The closed-face MTBM had an outer diameter of
1.472m, creating a theoretical overcut annulus of 19mm
around jacking pipes of outer diameter 1.434m. As mentioned
earlier, the MBTM used the slurry transport system for
maintaining a stable tunnel face, as well as for thetransportation of excavated spoils. A shield embodied themachinery in the MTBM, with a succeeding sleeve attached.
The cutter face was outfitted with four roller cutters arrangedon the face, and three roller cutters at the peripheral forformation of the overcut.
Site data of measured jacking loads (Figure 4) from the earlier
described site was used for benchmarking against predicted
frictional forces. Site data could be generalised under two
drive sections. The jacking forces within the initial drive
portion (Section 1 – 0 to 60m) accumulated within a lower
bound of 7kN/m and an upper bound of 12kN/m. The latter
drive portion (Section 2 – 60m to 108m) exhibited gains infrictional loads of between 71kN/m (upper bound) and
41kN/m (lower bound). The positions where the samples
were collected from are entailed in Figure 4 below, including
the respective shear box test results.

853

Sample 1
ϕpeak = 36°
cpeak = 49kPa
ϕres = 28°cres = 37kPa

Sample 2
ϕpeak = 49°
cpeak = 78kPa
ϕres = 47°
cres = 4kPa

Figure 4 Back-analysed upper and lower bound jacking
forces with respect to measured values including effect of
stoppages in jacking in Section 2 only
Using Eqs. (1), (2) & (3) together with the residual strength
properties of the excavated spoils, the required frictionalcoefficients for back-analysis were determined and presented
in Figure 4.
The results were based on the Terzaghi set of parameters from
Table 2. For the initial portion of the drive, the back-analysed
frictional coefficients were 0.19 (lower-bound) and 0.21
(upper-bound), while those for the latter portion of the drive
were 0.30 (lower-bound) and 0.42 (upper-bound). Withreference to Table 3, the lower-bound coefficients of friction
lie within the range of values reported for lubricated drives.
The upper-bound frictional coefficient lies beyond the rangeof acceptable values for lubricated drives.

Figure 5 Back-analysed upper and lower bound jacking
forces with respect to measured values, but with effect of
stoppages excluded from Section 2 only
resistance and effects of lubrication are usually mitigated at
this initial stage of work. This is confirmed by the
back-analysed coefficients, which are comparable to those forunlubricated sliding friction for concrete on gravel or sand as
per Table 3.

3.3 Effect of stoppage on jacking forces
In Section 2 of the drive, the disparity between upper andlower bounds for the coefficients could be explained by the
stoppages in jacking works. Jacking forces in this latter
section have been segregated as shown in Figure 5, thus
Figure 5 shows jacking forces associated with only on-goingjacking works. Frictional coefficients for the lubricatedSection 2 are seen to have reduced to 0.32 (lower-bound) and
0.34 (upper-bound). These coefficient values were slightly
above the upper limit of the corresponding values reported inTable 3 (under lubrication component). Figure 6 shows
jacking forces, with forces attributed to stoppages presented
for Section 2 of the drive. Frictional coefficients for Section 2
are 0.36 (lower-bound) and 0.42 (upper-bound), which are
generally higher as re-start of jacking works after periods ofstoppages are known to induce greater static frictional

Figure 6 Back-analysed upper and lower bound jacking
forces with respect to measured values, with only effect of
stoppages in Section 2 only854

Concrete on
gravel or sand(sliding)

pipe-jacking process as had been described in detail earlier
and interpreted from Figures 4, 5 and 6.
Figure 8 illustrates the occurrences of extended stoppages,
where sudden increments in jacking forces have beenpreviously presented. These were not weekend stoppages as
jacking was on-going through the weekends. Such stoppages
in work could have been due to major maintenance works,
possibly from the replacement of roller cutters. During suchstoppages, lubricant injected into the overcut around the pipe
peripheral could be lost through fissures in the surrounding
geology as shown schematically in Figure 9.

Concrete on
gravel or sand(static)

Section 2 (during jacking)μ = 0.34 (upper)
Section 2 (stoppages)
μ = 0.42 (upper)

Section 2 (during jacking)μ = 0.32 (lower)

Section 2 (stoppages)
μ = 0.36 (lower)

Section 1 (on-going)
μ = 0.21 (upper)
Section 1 (on-going)
μ = 0.19 (lower)

0

0.1

0.2

0.3

0.4

0.5

0.6

Partial lubricationin overcut due tofissures

Loss of lubricant
into fissuresμJackingpipe

Full lubrication in
overcut

Fluid friction(lubrication)
Concrete-clay(sliding)

Concrete-clay(static)

Figure 7 'Dynamic' behaviour of coefficient of friction
during various stages of the pipe jacking process

Figure 9 Schematic idealisation showing
lubrication through fissures in a rock massloss ofThe pressure for lubricant injection would also be relieved,
mitigating any buoyancy effect that may have reduced thefrictional forces. The effect of loss of buoyancy supporting
the weight of the pipe could result in higher pipe-soil contact
area. There was an observed change of the lubricating regime
in Section 2 of the drive, particularly upon the advent of thethird major stoppage. This increase in the volume of lubricant(from 170 litres to 800 litres per metre run) was in reaction to
the increased rate of frictional loads upon transition from
Section 1 to Section 2 of the drive. However, this intensified
effort had only been initiated after several spikes in thejacking load had been accrued.
4. CONCLUSIONS

Figure 8 Development of jacking force and elapsed timein relation to stoppages and volume of lubricant injected
into tunnel overcut
Figure 7 shows the ‘dynamic’ propagation of friction
coefficients during different stages of work in theThis paper has highlighted the potential of using direct shear
testing of excavation spoils to obtain residual strength valuesof geologically ‘young’ weathered rock to predict jacking
forces. The main advantage of using reconstituted excavation
spoils as specimens is the ability to obtain rock strengthparameters which otherwise would be elusive as intact andcompetent rock cores (≥100mm in length) are usuallydifficult to obtain in such weathered and friable conditions.
One of the main factors affecting the accuracy of the
back-analysis of the lower and upper bound jacking forces isthe ‘dynamic’ or fluctuation nature of the coefficients offriction as the pipes penetrate into differing geology within asimilar drive and also due to its construction activity (whether
jacking is on-going or stopped for maintenance). The
back-analysed coefficient of friction seems to be also affectedby the fact that whether the drive is fully or partially855

lubricated. Hence, soil-structure interaction is indeed an
important consideration in any pipe-jacking work.
Jacking forces have been observed to have increased during
the re-start of jacking activities after periods of stoppages as
more thrust is required to overcome static frictional resistanceprior to the realisation of the positive benefits of lubrication.
However, it is the intention of the Authors to carry out moretests to further verify the reliability of this method for a wider
variety of local rock types, especially in geologically ‘young’
weathered rock.
5. ACKNOWLEDGMENTS
The authors would like to thank the generosity of Hock Seng
Lee Bhd and Jurutera Jasa (Sarawak) Sdn Bhd for theirsupport during this study. The authors are also thankful to theground staff of Shaft & Tunnel Sdn Bhd for their assistanceand patience during the study of the microtunnelling drive.
The work described in this paper would not have beenpossible without the efforts of Leong Hsiao Yun and Lionel
Foo Fang Ho.
6. REFERENCES[1] Norris, P. and G.W.E. Milligan, Field instrumentation
for monitoring the performance of jacked concrete pipes,
in 3rd International Symposium on Field Measurementsin Geomechanics, G. Sørum, Editor 1991, A.A.
Balkema: Oslo, Norway.
[2] Milligan, G.W.E. and M.A. Marshall. The functions andeffects of lubrication in pipe jacking. in World Tunnel
Congress. 1998. Sao Paulo, Brazil: Balkema.

[3] Milligan, G.W.E. and M.A. Marshall. The influence of
lubrication on jacking loads from six monitored pipe
jacks. in 16th International ISTT No-Dig 1998. 1998.
Lausanne, Switzerland: ISTT.
[4] Pellet-Beaucour, A.L. and R. Kastner, Experimental andanalytical study of friction forces during microtunneling
operations. Tunnelling and Underground SpaceTechnology, 2002. 17(1): p. 83-97.
[5] Staheli, K., Jacking force prediction an interface friction
approach based on pipe surface roughness, in School of
Civil and Environmental Engineering2006, Georgia
Institute of Technology: Georgia. p. 385.
[6] Shou, K., J. Yen, and M. Liu, On the frictional propertyof lubricants and its impact on jacking force andsoil–pipe interaction of pipe-jacking. Tunnelling andUnderground Space Technology, 2010. 25(4): p.
469-477.
[7] The Australian Standards, AS1289.6.2.2-1998 Soilstrength and consolidation tests - Determination of theshear strength of a soil - Direct shear test using a shear
box, 1998, Standards Australia: Australia.
[8] Bagherzadeh-Khalkhali, A. and A.A. Mirghasemi,
Numerical and experimental direct shear tests forcoarse-grained soils. Particuology, 2009. 7(1): p. 83-91.
[9] Stein, D., K. Möllers, and R. Bielecki, Microtunnelling :Installation and Renewal of Nonman-Size Supply and
Sewage Lines by the Trenchless Construction Method.
1989, Berlin, Germany: Ernst.
[10] Tan, D.N.K., Geology of the Kuching Area, West
Sarawak, Malaysia, 1993, Geological Survey of
Malaysia: Kuching.

856

An idea of multi-functional storage reservoirs in mountain regions

Jan Jaremski
Rzeszow University of Technology, Rzeszów , Poland, jjaremsk@prz.edu.pl

KEYWORDS: flysch eluvium, reservoir, rock eluvium , mud avalanches , swelling
ABSTRACT: The paper presents possibilities of construction of multi-functional reservoir in mountainregions, especially those with soft rocks. Construction of relatively small reservoir instead of big ones andtaking the area configuration into account allows to eliminate expensive embankments and insists toconstruction of cofferdams. The cofferdam can be formed of the occurring rock eluvium. Such idea wasanalyzed for the south of Poland. At present we can observe the climate changes accompanied by increase of
the weather dynamics and unusual intensity of extreme phenomena. Mineralogical composition of rocks ortheir eluvia occurring in the considered area is usually not taken into account while forecasting of suchphenomena. The proposed tests on determination of the eluvium areas especially susceptible to formation of
mud avalanches in the mountains sites.
geological structure, i.e the laminar Carpathianflysch. It is water-permeable (there are no faults
and dislocations), so in this region there are very
good conditions for small flood pools. Such
conditions allow for high surface flows which
can be significantly slowed down owing torenaturization of the flows and so-called “small
retention”. The program of “small retention”
gives the best results in mountain and forest areas,for example the Bieszczady mountains in thesouth of Poland. “Small retention” satisfies mostrequirements connected with ecological methodsof water economics, and if it is correctly realized,
it can help in improvement of natural valuesof ecosystems []. The Podkarpacie region
belongs to the mountains range of the Carpathianarc and the fall before the Carpathians. The region
is very diversified considering its configurations
and geological structure. Such diversificationinfluences hydrography, climate, soils, flora andfauna etc.
Intense precipitations in the summer are very
important for the idea of construction of flood
pools. The mountain part of the region belongs to
the Carpathian climate with thermal conditions
varying depending on the site height.
Geologicalandhydrogeologicalconditions of the region in its mountain part arereally suitable for the flood pools. The eastern
Carpathians are built of flysch formations,weathered in their top layer. The eluvia usuallyoccur as dust or dust clays. The external partof the massif contains flysch formations (Fig. 1),
i.e. sandstone, mudstone, menillites, etc.[14].

1. INTRODUCTION
The climate variation has been observed recently.
They are accompanied by increase of the weather
dynamics and unusual intensity of extreme
phenomena. Precipitations are very intense andthey are accompanied by landslides and mud
avalanches, or long-lasting droughts. A lack ofsuitable retention causes a fast flow of discharge
water to the seas. Mineralogical composition ofrocks or their eluvia occurring in the considered
area is usually not taken into account while
forecasting of such phenomena.
Colloidal
activity, a state of colloidal bonding andcrystallization strongly influence formation ofavalanches,
strength of the mixtures andreversibility of their reactions with participationof clay minerals.
Polish water resources are rather poorin comparison with other European countries.
1600 m3 falls to one person in a typical year; in a
dry year it is below 1450 m3 [1]. In Spain such
number is twice higher, and for all Europe thisnumber equals to 4500 m3. Retention ability ofPolish reservoirs equals to 6% of the mean flowof the surface water per year. Moreover, it does
not prevent against periodical excess of deficitof water. Our water resources vary at different
seasons and their distribution is not uniform. The
Podkarpacie region, especially its mountain part,
belongs to the regions where the water resources
are the most poor.
Only 20% of all
the underground water of that region occur in its
mountain part. In this region there is a lack ofrich aquiferous layers because of an unique857

Fig. 1. Outcrops of the Carpathian Flysch
The Bieszczady Mountains are the area where
there are very good conditions for the considered
reservoirs because of favourable tectonics and its
effects. It appears from the tests realized in the
Podkarpacie region it appears that the eluvia
occurring under the surface are of a high volume,and from mineralogical tests it results that they
are suitable for location of reservoirs because theyguarantee the maximum tightness. Thus,
construction of small earth dams or rock-filldams probably will not require application of
additional sealing elements.
The eluviumof suitable parameters will play that role. The
eluvia in the layers located just under the surface
usually occur as dust or dust clays.
Generally speaking, in the flysch formationsof the Podkarpacie region there are no advanced
aquiferous layers. The minimum surface
permeability is also an effect of mineralogicalstructure of the surface layers. Under moderate
swelling, illite minerals seal the site surface andincrease intensity of the surface flow. In a
consequence, even relatively low precipitations
cause a flood situation at many water-courses
and rivers in the Poskarpacie region. Thus,
construction of multi-functional pools of small
retention on the limited areas protects the watercourses of the Podkarpacie region against
the results of floods. The proposed idea increases
pure water resources, influences beauty of theregion. It also changes geotechnical conditions
and improves stabilization of slopes, protects
against landslides.
The proposed idea seems to be reasonablebecause in 1995 in Japan 360 such reservoirs
were built in the mountains, and the country wasprotected against floods. At the same time,
suitable amount of water was delivered for rice
growing (Japan is an important exporter of riceat present).
Let us note that the considered pools are locatedin the mountains, and the cofferdams baffling thevalleys are made of soil concrete becauseof a large amount of montmorillonite in the rock
eluvia.
In the past in Andalusia the Moor
engineers built the reservoirs used for the
purposes of agriculture, and this area of Spaindeveloped very quickly. Let us remember that the
proposed pools should be supplied by waterflowing from the eastern Carpathians being an
enclave of unique environmental conditions.
Water-bearing
character of theflysch rocks isconnected with
their slots –it results from
their
low
porosity,

i.e. from 0.02 to 0.1. The flow in the interporous
spaces is difficult because of their small
dimensions; moreover they are filled withthe weathered illite fraction, susceptible toshrinkage and swelling, decreasing water
absorbability of the flysch, sometimes to zero.
Water absorbability gets worse as the depth oflayer deposits rises. Resources of undergroundwater is very low, there are no water-bearing
layers. Water is stopped and occurs only in thecrackedrocks. Under
such conditions,precipitation waters can easily flow to the watercourses, and next to the fall before the
Carpathians where they supply the reservoir ofunderground waters. From that reservoir theyflow through the underground gutters to the basinof the Dniestr River. Minimal surface
permeability is also an effect of mineralogicalstructure of surface layers, i.e. illite and smectite
minerals. Because of swelling, these minerals seal
the surface and increase intensity of the surface
run-off. In a consequence, even relatively small
amount of precipitations cause floods at many
rivers and water-courses of the Podkarpacieregion.
Climate variation causes serious rain-storms
which are very dangerous for inhabitants of the
Podkarpacie region. Such storm took place on the
26th July, 2005 near locality Zagórz. Some
villages were flooded in a very short time.
The mentioned above example of Japan also
proves that the proposed solution seems to bereasonable.
2. THE PREDICTED KINDS OF
POOLS

FLOOD

Rational management of water resources in the
Podkarpacie region requires minimization of theirvariation at time and space (periods of drought
and flood). Thus, retention must be developed in
the area of the basin. Construction of small multifunctional pools is recommended. In managementof each pool one function here is usually
dominating (flood protection, touristic, ecologicalor energy function and so on), and other tasks are
assigned as that main function. In practice, there
are pools satisfying two or three main economic
tasks at the same time depending on local needs. .
There is no one official classification offlood pools. So-called “small retention” includes
all the microreservoirs, and small pools positively
influencing the environment [10]. Small pools areusually of total volume to 0.1 of million
m3 and the maximum swell height 5 m. They arebuilt one by one, or in a cascade system [12]. All
the cascade is usually used for energetic purposes.
The systems of small retention pools are proposed
instead of one big reservoir. In Bieszczady
mountains there are many river valleys (Fig.2)
which are good places for the proposed flood858

pools. In the past many people lived in thosevalleys, at present they uninhabited.

3. APPLICATION OF FLYSCH ELUVIAFORCONSTRUCTION
OF
COFFERDAMS

According to the proposed concept, reservoirs –depending on their location – can be flood pools,
equalizing tanks, they can also be auxiliary
reservoirs near big flood pools[3]. Some
reservoirs should play the role of flow reservoirs
formed as a result of partition of water-courses.
Dry reservoirs are also provided as anti-flood
protection. They will be accumulate the flowing
eluvium or organic products.

From the investigations on landslides in the
Podkarpacie area it appears that the eluvia
existing under the site surface are of a large
thickness. From mineralogical tests it results thatsuch eluvia are very good for locationof reservoirs because they provide the maximum
tightness. Thus, construction of small earth dams
will not require additional sealing elements. Thisrole should be played by a layer of eluviumof suitable parameters.
Experience obtained while investigations
on marl eluvia as well as the elaborated test
methods were applied for investigationsof he Carpathian flysch and dust soils occurring in
the river terraces, grounds of loess massif, illite
clays and baidelite clays [4,5]. These grounds are
susceptible to moisture changes accompaniedby shrinkage and swelling as well as petrification.
This property is strictly connected with presenceofillitein mineralogicalcompositionof the considered media. Even a very small
amount of that mineral influences joints of thebasic minerals.
The samples prepared of eluvium shales
were tested and next the obtained results werecompared with the results obtained for the
samples from river valleys treated as theaccumulated sediment of the flysch eluvium [5].
The test were carried out in order to estimate
usability of flysch eluvia for construction ofcofferdams by identification of clayey minerals.
Differentiation of these minerals is importantbecause the contents of smectite groups exerts anunfavourable influence on setting and hardeningof ground concrete [11]. As for the consideredground media, their basic geotechnical parameters
were tested and special attention was paid toswelling tests allowing to distinguish clayey
minerals. Before the swelling tests, physical,
chemical and strength tests as well as
mineralogical tests of the considered media werecarried out [4, 6]. Recognition of the grounds was
basedonchemicalandmineralogicalcompositions by means of fluorescence, X-ray
difractommetry and scanning microscopy.
Moreover, derivatographic analysis was realized.
Recognition of microstructure of the clays with
so-called hidden swelling parameter were carriedout by means of scanning microscopy. The
mentioned tests of chemical composition (Fig.3)
show the presence of potassium which occurs
only in illite. It was proved by tests realized
by means of the scanning electron microscope (SEM).
The tests of flysch swelling, swelling clay or the
media with so-called hidden swelling parameterused the method described previously [8]. This

Fig. 2. An exemplary site for a reservoir in the
Bieszczady mountains
In the considered region, the intense
precipitations, of big slopes of the ground and theflysch substrate cause intense run-off of theweathered flysch layer, sometimes being like anavalanche. The rubble occurring in the reservoirs
and flumes of water-courses causes that they
become more and more shallow.
Thus,
construction of dry reservoirs is. In the considered
area there are unusual conditions of flysch
weathering, especially biological weathering.
Such conditions are a result of rich afforestation.
Generally speaking, in the considered region plan
dry reservoirs it will be possible to plan dry
reservoirs where will be accumulate the flowing
weathered eluvia or organic products. Such places
will be the programmed enclave of the intensedevelopment of flora and fauna and, we can also
expect favourable humus processes.

859

mineralogical composition should be taken
into account. Tests of susceptibility of the softrocks for formation of the mud avalanches could
be limited to drawing of eluvium samples
occurring at the slope. The samples should be
subjected to swelling tests by means of themethod given in the some previous papers[4,8,11]. This simple method allows to identify
clayey minerals responsible for swelling,occurring in the rocks. While tests we use thepresence of endothermic maxima – thetemperature points at which the heated sample
loses contents of bound water. The author foundthat minerals from the groups of monmorillonite
and illite (causing swelling) lost that waterat the temperature
up
to
200
degrees,and kaolinite – at 600 degree. Thus,
the comminuted eluvium sample prepared to thestandard swelling test and free from water,
subjected to the swelling test manifests swellingup to tens %. Under such conditions, the kaolinite
sample swells up to some %. Let us rememberthat montmorillonites soak to 700%, while
kaolinite soaks to 90%.

method allows for cheap tests of flysch eluvia
and estimation of their usability. Identification ofillite in the flysch eluvium allows to qualify it asa material for construction of cofferdams becauseof moderate shrinkage and swelling as well as
favourable petrification and sealing properties.
Moreover, the proposed tests could help inselection of places for reservoir location into
taking a proper tightness of the ground account.
Presenceofillitebeing
an aluminosilicate
responsible
for eluvium
petrification allows for application of flysch chips
and flysch eluvia for construction of earth
cofferdams
sccordingtotheproposed
construction of multi-functional pools in the
Bieszczady mountains [3]. In Japan, the flood
pools were located in the mountains and thecofferdams in the valleys
were the ground-concrete is made. The eluvia
occurring in Japan contain montmorillonites, so
formation of cofferdrams is difficult, because
it is necessary to add much cement to the eluvia.
It is also necessary to take care of groundconcrete because of formation of scratches causedby shrinkage of montmorillonite [13].
Construction of cofferdrams using flysch
chips and flysch eluvia is considered for eachreservoir in Bieszczady, depending on its
geological structure. Of course, a suitable amountof clayey minerals for petrification and sealing isnecessary. If there are no enough illite minerals
in the eluvium, we must consider a solution using
geotextiles and sealing foils for cofferdrams, andmaterials offered by such firms as Tensar
International
Corporation,
Ten
Cate
Geosynthetics and Geobrugg.

4.1 THE PROPOSED METHODEXAMINATION OF CLAY MINERALSOF

Application of the method of swelling ability test,
given in the author’s earlier works [2,4,8] wasproposed as method of determining the content of
illite and smectite minerals. It has been prooved
so far that it is just these minerals thatare responsible for moisture changes, extreme
strength parameters and processes of petrificationand maximum moisturizing.
Tests were carried out on the material taken fromthe deposit of different breakups, and samples
were grouped in accordance with the diameter ofsieve meshes. With the method of successive
trials, it has been found that the most stable andreal results are obtained by conducting the tests inthe following way [11]:4.
THE
PROPOSEDTESTSFORDETERMINATION OF THE ELUVIUM
AREAS ESPECIALLY SUSCEPTIBLE TO
FORMATION OF MUD AVALANCHES IN
THE MOUNTAINS SITES
Experience obtained in the while works on marleluvia [2,4,7], and the elaborated test methods
were used during tests of Carpathian flysch anddust soils occurring in river terraces generated
as a result of the eluvia accumulation [5,11].

1.

2.

The samples prepared from the shale eluvia were
tested and compared with the samples from theriver valleys, treated as the accumulated sedimentof the flysch eluvium. The basic geotechnicalparameters of the considered ground media were
tested. Special attention was paid to swelling tests
allowing to identify clayey minerals. At present,
the author is engaged in works connected withthe causes of mud avalanches formation. In suchcases, hydraulic conditions and influence of

3.
4.

860

the material undergoes a preliminaryairdrying in the case of water content
so that in can be sifted through the of
Ø=0.76 mm
the material thus prepared is subjected tomoisture content near to the plasticity
limit(during the trial of roller bursts at
the first or second instant) and issubjected to working ( just like a typical“dough” )forming the material into the “cake” –like shape of 10 mm in heightthen,sampleswith
thediameter Ø= 50mm and 65 mm and the

5.

6.
7.

8.

height of 10 mm and 20 mm are cut bymeans of a ring
the samples thus prepared are weighed
and dried in a drier at a temperatureof 200 C for the period from two hours
until reaching constant weightafter drying and cooling process
the samples are weighed to determine
their water content
thesamplesaresubjected
for the examination of swelling minerals
in the apparatus testing unboundedswelling
after swelling has been defined the final
moisture is determined.

utilization of SEM, derivatogram method,
fluorescence, diffraction method,.
Simple analyses and tests according to thediscussed swelling tests allow to identify sites
which can be easily subjected to mud avalanches
and rock falls. Such information are very usefulfor the site management.

The considered test informs us about the amounts
and kinds of swelling minerals in the point ofsampling. Thus, we get also the data concerning
ability of the tested rock medium to water
stoppage and moisture degree, as well as the
maximum moisture occurring under natural
conditions.

6.

CONCLUSIONS

6.1. Suitable geological (soft rocks) structure
allowsfor
cheap
constructionof reservoirs. Dislocations and fault
fissures filled with weathering materials
because of their mineralogical structure,i.e. swelling minerals cause self-sealingof the bottom. A lack of water-bearing
layers eliminates uncontrolled outlet ofwater to the ground.
6.2. Distinct improvement of protection
against floods is an advantage of the
proposed solution (see the mentioned
solution in Japan). At present, intense
precipitations are transported to watercourses and rivers because they cannot
percolate to the ground (it is favourable
for reservoir structure).
6.3. Water supply for inhabitants of a given
region and water collecting in reservoirs
can be very important in the future.
Large amount of water of high quality
should be a valuable good. Water
resources could influence future
development of many regions.
6.4. Constructionoftheconsidered
reservoirs is rather cheap because it ispossible to introduce unified earth
cofferdrams using ground-concrete as aconstruction material.
6.5. Introductionoftheconsidered
reservoirs could help in generation ofnewplacesof employment(construction
and service,
tourists’
service)
6.6. Some reservoirs could be usedin energetics.

Fig.3.Examplesof rocks samples after the investigationAfterswelling,thesamplemoistureis the maximum moisture, under which we get theworst numerical values of the strength parameters.
Under such conditions, landslides are possible. In
our analysis, soft rocks are treated as mixturescontaining much clayey minerals. In the case ofoccurrenceof greatercontents
ofmontmoryllonite, greater amount of moisture ispresent and it is accompanied by fluidizationand formation of mud avalanches. Similarly,
content of the swelling minerals in higher parts ofmountains causes their collecting in the rock slots.
Then, possibility of landslides and rock falls
increases.
The swelling investigation
were precededby physical, chemical and strength tests of theconsidered media. Recognition of such soils werebased on tests of chemical and mineralogicalcomposition by means of fluorescence and X-ray
diffraction method . Moreover, derivatographic
analyses were performed. Tests of clay structurewith so-called swelling parameter were realized
by meansofscanning
microscopy(SEM)

7.

861

REFERENS

[1] Bogacz – Rygas M : (2007) Aspekty
ekologiczne stosowania nowych rozwiązań
odprowadzania
wód
opadowychiichzagospodarowanie(in polish),. Wavin MetalplastBuk.Kielce.
[2] Jaremski J.:(1994) Influence of physical andchemical processes occurring in eluvium of theOpole marls and their influence on thegeotechnical parameters, Proceedings of the 7thInternational Congress IAEG, Lisboa , A.A
Balkema, 859-868.
[3] Jaremski J.:( 2007) O niektórych aspektach
budowywielofunkcyjnychzbiorników
retencyjnych w części górzystej województwa
podkarpackiego. Cywilizacja i żywioły, (inpolish), seria: Monografie Instytutu Meteorologii
i Gospodarki Wodnej, Warszawa.
[4] Jaremski J.:(1995) Proposal for a method
determining value of the maximum swelling ofsoil by example of the Opole marls eluvium andillite clays. Proc. of 8th Int. Cong. on RockMechanics, Tokyo, A.A.Balkema.
[5] Jaremski J.:(2004) The influence of changes
of weathering water content on the formation oflandslides in the Carpathian Flysch. Proc. of 9thAustralia New Zealand Conference onGeomechanics, Auckland – February 2004, New
Zealand Geotechnical Society Inc.
[6] Jaremski J,:(2005), The influence of physicaland chemical processes occurring in theweathered Opole marls on the geotechnicalparameters, Publishing House of Rzeszów
University of Technology, 2005,p. 98.
[7] Jaremski J.(1991), Creep property of Opole
marls weathering on the base of the model
laboratory research and the research “ in situ” ,Proceedings of the 7th International Congress on

Rocks Mechanics, Aachen, A.A. Balkema, 1991,
269-273.
[8] Jaremski J:(1994) Proposal for method ofstating the value of maximum swelling of the soilon the example of Opole marls eluvium. Proc.
3th Int. Conf. on Trop. Resid. Soils., Maseru ,A.A.Balkema,[9] Jaremski J:(1990) Analysis of swelling of marleluvium of water infiltration. Proc. 6th Int. Cong.
IAEG. Amsterdam, A.A. Balkema, Rotterdam .
[10] Jaremski J.:(2010) Aspekty geotechniczne w
ochronie zasobów wodnych województwa
podkarpackiego. Inżynieria Morska i Geotechnika(in polish), nr. 1. Gdańsk 2010.
[11] Jaremski J.:(2010), Influence of moisturechanges on physico - chemical processes andgeotechnical parameters in soft rocks , Studia
Geotechnica et Mechanica, Vol. XXXII, No. 1,
2010[12] Mioduszewski W.:(2008) Mała retencja w
lasach elementem kształtowania i ochrony
zasobów wodnych. Studia i Materiały Centrum
Edukacji Przyrodniczo-Leśnej R. 10. Zeszyt 2(18) (in polish) .
[13] WAKIZAKA Y.:(1994) Influence ofmontmorillonite group Bering aggregates on
concrete. 7th International IAEG Congress, A.A.
Balkema 1994, s. 3285-3294.
[14] Zabuski L., Thiel K., Bober L.(1999),
Landslides in Polish Carpathian Flysch, (inpolish), IBWN PAN, Gdańsk, p. 172.

862

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Estimation Method of Amount of Tsunami Disaster Wastes during the 2011 off
the Pacific Coast of Tohoku Earthquake
Minoru YAMANAKA1, Naoya TOYOTA2, and Shuichi HASEGAWA3, Atsuko NONOMURA4
1, 4
Associate professor, Kagawa University, Japan 2 Graduate student, Kagawa University, Japan
3
Professor, Kagawa University, Japan
ABSTRACT
A huge amount of disaster wastes was accumulated along coastline areas of the Tohoku region by tsunami during the2011 off the Pacific coast of Tohoku Earthquake. It is very important to estimate an amount of tsunami disaster wastes
as for quick recover and revival of damaged areas. This paper describes a method of wastes amount estimation based on
the officially announced data analyzed in GIS platform. It was found that there is a good relationship between theamount of the disaster wastes and land use, land cover and tsunami height. Therefore, it is necessary to subdivide theaffected area according to land use and cover, the tsunami height and so on.
Keywords: disaster wastes, tsunami, earthquake, estimation

1. Introduction
A huge amount of disaster wastes was accumulated along
coastline areas of the Tohoku region by tsunami during the2011 off the Pacific coast of Tohoku Earthquake. The total
amount of tsunami disaster wastes was estimated to be 26.7
million tons. It is 50 % of the total amount of the waste ofJapan in 2009.The Tohoku earthquake caused widespread
damage. Since the coastal line of Tohoku region is highly
indented, it is difficult to find secure places for piling thewastes. Moreover, road is only a means of transportation ofwastes to the waste disposal site. All port facilities werealready damaged. This means Japan is in extremely
difficult situation to dispose the disaster wastes.
Since magnitude or height of tsunami in Shikoku region
due to predicted earthquakes such as the Tokai Earthquake
and the Nankai Earthquake is comparable to the Tohoku
Earthquake, anticipated amount of disaster wastes in
Shikoku is also very high. It is very important to estimate an
amount of tsunami disaster wastes for quick recover andrevival of damaged areas.
This paper describes a method of wastes amountestimation based on the officially announced data analyzed
in GIS platform.
2. Characteristic and amount of disaster wastes
2.1 Characteristics of the disaster wastes
The characteristics of earthquake, flood and tsunami
disaster wastes are shown in Table.1. They are look alikeoccur suddenly and extensively. However in tsunami
disaster wastes, it is necessary to secure large places forpiling the wastes, since the huge amount of wastes
contained large volume of debris mainly discharged byheavy machinery. Similarly, tsunami disaster wastes
contain various kinds of materials, such as waste wood,
metals, salt water. In case of earth and sand mixed withdisaster wastes, unless the disaster wastes are not removed,
the earth and sand are also dealt as disaster wastes. In orderto decrease amount of the earth and sand in disaster wastes,
it is aimed to utilize a technology for separating earth andsand from the wastes. Wood waste can be recycled, butsalinity, earth and sand, metals can possesses harmful effectwhile recycling or incineration of wood.
2.2 Amount of disaster wastes of main previous
earthquake
The amount of disaster wastes and situation of damage due
to earthquake is shown in Table.2, and the status in
Rikuzentakata city damaged by the tsunami is shown in
Fig.1 (a)-(b). The amount of disaster wastes of Great
Hanshin Awaji Earthquake in1995 was about 20 million
tons [2]. About 640 thousand houses were damaged in theearthquake and the amount of wastes was about eight times
as much as an amount of wastes of Hyogo region on a year
in 2006. Fortunately, most of them can be disposed in
Phoenix Landfill Disposal Site near the damaged areas. The
amount of disaster wastes of the Niigata Chuetsu
Earthquake whose seismic centre is on the mountainous
area in 2004 was about 500 thousand tons. About 120
thousand houses were damaged in the earthquake. 92 % of
the disaster wastes were destruction wastes. Concrete
debris (51 %) and waste wood (25 %) accounted for 76 %of the total [2].
On the other hand, the amount of flood disaster wastes of
Typhoon No.16 in 2004, which struck Takamatsu city, was23,123 tons. It was twice as much as a month of wastes on
Takamatsu city in 2003. Since the wastes contained salinity,a large amount of unusable household goods wasabandoned. Therefore, it took long time to manage thedisaster wastes [3].
2.3 Unit amount of disaster wastes in published paper
Estimation of unit amount of disaster wastes also
corresponds amount of past disaster wastes per householdor per square kilometer.
Nakamichi estimates unit amount of the disaster wastes
to be 0.571 t/km2 on wooden constructions or 1.47 t/km2 on
RC buildings, and 1.27t/km2 on steel constructions. He did
this estimation from amount of the disaster wastes of Great863

Hanshin Awaji Earthquake [4].
Shimaoka estimates unit amount of the disaster wastes tobe 0.31 t/km2on wooden constructions or 1.09 t/km2 on RC
buildings, and 1.80 t/km2 on steel constructions [2].
Hirayama estimates unit amount of the disaster wastes
considering the extent of household’s damage on flood
disaster to be 12.9 tons per household on completely
destroyed, 6.5 tons per household on a half destroyed, 2.5
tons per household on partially destroyed, 4.6 tons perhousehold on inundation above floor level, 0.62 ton perhousehold on inundation under floor level [5].The unit ofdisaster wastes for estimating amount of the tsunami
disaster wastes in Tohoku earthquake was 61.9 tons perhousehold or 113 tons per a building [6].
3. Amount of tsunami disaster wastes
The amount of disaster wastes in each municipal
corporation, damaged by tsunami, are shown in Fig.2 (a)~(c). Fig.2 (a) illustrates the amount of disaster wastes ismuch in Ishinomaki and Higashimatsusima, Sendai, which
are highly built-up areas. Fig.2 (b) shows the amountdisaster waste per flooded areas is high in Kamaishi andRikuzentakata where were damaged by more than 10 m oftsunami height. In Fig.2 (c), Yamamoto where flooded area
is relatively small.

Table.1 Characteristic of the each disaster [1]typeoccurrences

Characteristics

Earthquake disaster

Flood disaster

Tsunami disaster・Occur heavily and suddenly
・Low earthquake-resisting
capacity buildings weredamaged・broken up by a crane・Occur heavily and suddenly
・The lower – lying districts aredamaged by the collapse of theembankment.
・The proportion of householdrefuse is high.

・Occur heavily and suddenly
・It is necessary to securelarge places for leaving thewastes.

・composed of household anddebris of fallen houses
・There are many recyclable,
reusable things among thewastes.
・It can be expected to separate
wastes by demolition
management.

・household wastes byinundation above or under floor
level.
・It is difficult to separate on thesite.
・The wastes are perishable
since it contains plenty of waterand sand・The wastes contain various
kinds of materials, such aswaste wood, metals, salt water.
・Salinity contained thewastes sometimes have a
harmful effect while recycling
or incineration.

Table.2 Amount of disaster wastes and situation of damage of main previous earthquake[2]NameGreat Hanshin Awaji
Earthquake

Niigata Chuetsu
Earthquake

Noto Peninsula
Earthquake

Niigataken Chuetsu-oki
Earthquake

Date ofoccurring

Jan.17.1995

Oct.23.2004

Mar.25.2007

Jul.16.2007

Magnitude

7.3

6.8

6.9

6.8

Damagedareas

Kobe, Nishinomiya,
Awaji-shima

Nagaoka, Yamakoshi,
Ojiya

Wajima, Nanao, Siga

Kashiwazaki, Kariwa

Situation ofdamageHouse:
Completely destroyed
104,906 houses
Harf destroyed
144,274 houses
Partially destroyed
390,506 houses

House:
Completely destroyed
3,157 houses
Harf destroyed
13,808 houses
Partially destroyed
103,854 houses

House:
Completely destroyed
638 houses
Harf destroyed
1,563 houses
Partially destroyed
13,553 houses

House:
Completely destroyed
1,224 houses
Harf destroyed
5,241 houses
Partially destroyed
34,277 houses

Amount ofdisaster wastes

About 20 million tons

About 0.49 million tons

About 0.43 million tons

About 0.36 million tons

Regions
Hyogo Pref.:Niigata Pref.:wastes in 2006 About 0.25 million tons About 1.12 million tons

Ishikawa Pref.:About 0.5 million tons

Niigata Pref.:About 1.12 million tons

864

140

Iwaki
Hironoamount of disaster wastes per flooded
areas (thousand tons/ km2)

HironoNarahaTomiokaOkuma
FutabaNamie
Minamisouma
ShinchiSouma

Watari

Yamamoto

Natori

Iwanuma

Tagajou

SendaiHirono

Iwaki

TomiokaNarahaFutaba

Okuma

Namie

Minamisouma

ShinchiSouma

Watari

Yamamoto

Natori

Iwanuma

Tagajou

Sendai

Shiogama

Shichigahama

Matsusima

Higashimatsusi…

Rihu

Onagawa

Ishinomaki

Minamisanriku

Kesennuma

OhunatoRikuzentakata

Otsuchi

Kamaishi

MiyakoYamadaIwaizumi

Fudai

Tanohata

Kuji

Noda

Hirono(c) Amount of disaster wastes per household of flooded areas
Fig.2 Difference of amount of disaster wastes by damaged municipal corporation in the Tohoku region

Iwaki
NarahaTomiokaOkuma
FutabaNamie
Minamisouma
Souma
Shinchi
Watari

Yamamoto

Natori

Iwanuma

Tagajou

Shichigahama
Shiogama
Matsusima

30

33

18

SendaiShichigahama
Shiogama

Rihu
Higashimatsusima
Ishinomaki
Onagawa
Minamisanriku
Kesennuma
Rikuzentakata
OhunatoKamaishi
Otsuchi
YamadaMiyakoIwaizumi
Fudai

Tanohata
Noda
Kuji
Hirono865

Matsusima
Rihu
Higashimatsusima
Ishinomaki
Onagawa
Minamisanriku
Kesennuma
OhunatoRikuzentakata

Otsuchi

Kamaishi
YamadaMiyakoIwaizumi
Fudai

Tanohata
Noda
Kuji
2000

12 29 47 25 36

amount of disaster wastes (thousand tons)

Hirono

0

62

46

29

21

81

81 85

67

81

95

74

89

121

106

83 91

103

109

130

132

105

102

100

132 140150

172 177

167

157

146

165

200

24
32
34
22
40

253

250

50

72
94
82
73
81

246

58
94

125
127
105
122
95

183
176

260
286

178
189
154

114

213
180 175
212
239
242
271

Fukushima
Miyagi(a) Amount of disaster wastes
Iwate

508

178
250
94
260 218129
17 81
11 77 57
176

700
640
738
838
820 756

55

37

50

410

494

500
450
400
350
300
250
200
150
100
50
0

4,458

2500

327

500

1352
148214351500

526
542 483732

1000

365 286
20 95

0

265

300

amount of disaster wastes per householdof flooded areas(thousand tons/ km2)(a) photo on April 6. 2012 .
(b) photo on June 5. 2012.
Fig.1 Disaster wastes in Rikuzentakata City(b) Amount of disaster wastes per flooded areas

5000

5000
4500

R² = 0.8894

4000

amount of disaster wastes(thousand tons)

amount of disaster wastes(thousand tons)

4500
3500
3000
2500
2000
1500
1000

3500
3000
2500
2000
1500
1000

500
0

R² = 0.8744

4000

500

0

10000 20000 30000 40000
households of flooded areas

0

50000

0

5

10

15

20

25

flooded areas (km2)(a) Flooded area(b) Number of household
Fig.3 Relationship between amount of disaster and each disaster data in Tohoku earth quake

4. Height of tsunami and unit amount of disaster wastes
It is considered the amount of disaster wastes in flooded
areas deals with the height of tsunami. The relationshipbetween the height of tsunami in each municipal
corporation and the amount of disaster wastes per flooded
areas is shown in Fig.4. There are two lines on therelationship because there is probably the difference inproportion of residential area in flooded area. In this case,further consideration is needed.
5. Conclusion
In conclusion, it is found that there is a correlation between
unit amount of tsunami disaster wastes and each disasterdata. It is now possible to contribute in Tonankai or Nankai
earthquake related possible disaster wastes. The following
are main conclusion of this study;1) We can estimate the necessary action for disaster waste
disposal or recommend suitable places for deposit thewaste.
2) We can estimate the scale of disposal site and the numberof roads or harbors needed.
3) Now estimation of amount of recyclable disaster wastes
is possible. Similarly, service plan for rebuilding andrecovery is also possible.
Reference[1]YAMANAKA Minoru, HATIMURA Tomoaki,
Problems about Disposal of Disaster Waste, Chishitsu to
chousa, pp.34-39, 2011[2]SHIMAOKA Takayuki, “Abnormal Weather andDisaster Waste accompanied by Global Warming”
Disaster Waste, Japan Society of Material Cycles andamount of disaster wastes per flooded areas(thousand tons/ km2)

Relationship between amount of disaster and each disasterdata are shown in Fig.3 (a) and (b). Fig.3 (a) shows that
relationship between the flooded area by the tsunami andthe amount of the disaster wastes has a positive correlation.
In addition, since it seems that the correlation is separated
by the magnitude of the flood area, it is needed to separate
the urban and agricultural land. And the correlation
between the number of households in flooded area and theamount of tsunami disaster wastes was found to be high asshown in Fig.3 (b). From this, unit amount of disaster
wastes as per household is estimated it 104 ton.

600
500
400
300
200
100
0

0

5

10

15

20

25

height of Tsunami (m)Fig.4 Relationship between the height of tsunami and theamount of disaster wastes per flooded areas
Waste Management Series,Chuohoki,2009,pp.3-20.
[3]Waste Management and Recycling Department:
Measure Guideline of flood waste, Ministry of theEnvironment, 2005.9.1[4]NAKAMICHI Tamehiro, “Waste Disposal Act of Great
Hanshin Awaji Earthquake” Disaster Waste, Japan Societyof Material Cycles and Waste Management Series Vol.3,
Chuohoki,2009, pp.23-46.
[5]HIRAYAMA Nagahisa, Estimation Procedure forQuantity of Disaster Waste from the Viewpoint ofEmergency Responses of Public Authorities at Flood
Disaster, J Environment system, Vpl.33, pp.29-36,2005[6]National Institute for Environmental Studies, About
Unit Amount of disaster waste Vol.1, Disaster Response
Network 2011.6.28[7]Statistics Bureau, Statistical Data Pertaining to the Great
East Japan Earthquake, Ministry of Internal Affairs andCommunications, 2011.11[8]HARAGUCHI Tsuyoshi and IWAMATSU Akira,Detailed Maps of the Impacts of the 2011 Japan Tsunami
Vol.1: Aomori, Iwate and Miyagi prefectures. Japan:
Kokon-Shoin,2011.
[9]HARAGUCHI Tsuyoshi and IWAMATSU Akira,Detailed Maps of the Impacts of the 2011 Japan Tsunami
Vol.2: Fukushima, Ibaraki and Chiba prefectures. Japan:
Kokon-Shoin,2011.

866

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Assessment of Runoff in the High Humid Foot-hill Areas of Arunachal
Himalayas Using Thornthwaite Equation
Md. Eahya Al Huda and Surendra Singh
Department of Geography, North-Eastern Hill University, Shillong-793022, India

ABSTRACT: Present paper examines daily runoff trend in Dikrong river catchment which is located in the lower
foot-hills of Arunachal Himalayas. The river catchment is characterized by the monsoon dominated hydrologic regime.
For the purpose, daily statistics of temperature, rainfall and discharge were collected from the various sources of state
Government of Assam, Guwahati and Government of Arunachal Pradesh, Itanagar and used for testing the validity ofThornthwaite-Mather [T-M] method considering Dikrong river catchment of high humid conditions. It is found that thismethod does not give the significant result for the correct assessment of daily Potential Evapotranspiration [PET] as
well as daily runoff [RO]. The fluctuation in predicted RO is found more than the observed one with it’s over estimation
during high rainfall intensity and under estimation in the period of low rainfall regime.
Keywords: Runoff, Potential Evapotranspiration, Thornthwaite-Mather method, Arunachal Himalayas, discharge rate

1. INTRODUCTION
There are numerous studies on the assessment of runoff in ariver catchment using universally accepted Thornthwaite
equation for the quick and correct prediction of runoff [1],[2], because this method is implicitly based on water balance
equation and simple to use with universally available
statistics of two hydro-meteorological parameters, namely,
precipitation and temperature.
In India especially in the most-humid areas where ‘saturation
excess’ processes of runoff generation are more prevalentrather than ‘infiltration’ processes in the monsoon dominatedhydrologic regime [3], there is a need of testing the validity ofThornthwaite-Mather [T-M] procedure to assess the runoff insuch high rainfall conditions. There are other methods thanT-M procedure used for the assessment of reference ET not
only in Indian conditions but across the world. In this regard,
Indian references are important to compare the suitability ofthese methods. For example, Mishra, et al. [4] used T-M
method for tropical regions of the country. Meshram, et al. [5]used other methods and found suitability of FAO
Penman-Monteith method, while Lenka, et al. [6] tested thevalidity of different methods like Blaney-Criddle [7], Hamon
method [8], Hargreaves and Samani [9], Papadakis [10], FAO
Penman-Monteith [11] that have different bases of PET
calculation for the prediction of reference ET. It is argued andfound that T-M procedure of the assessment of PET is simpler
than the other methods because it is water balance based
equation that is more appropriate to analyze the hydrologic
cycle while the other methods are based on atmosphericenergy parameters. Such other methods may be useful for the
study of meteorological phenomena rather than hydrologicalregime of a river catchment.
For the purpose, present paper examines the daily runoff
trend and its associated factors and also evaluates thesignificance of the application of T-M procedure taking into
account Dikrong river catchment of about 1,556 sq km area of

which more that 80% is under the hill topography with anaverage annual rainfall of 3,294mm.
1.1 Study Area
Dikrong river catchment is located in the foot-hills of
Arunachal Himalayas between 26˚55΄ to 27˚22΄N latitudes
and 93˚13΄E to 94˚0΄E longitudes with its transitional
characteristics of its location as it falls under Inter Tropical
Convergence Zone [ITCZ] where climate is monsoonal inthis part of Asian region [Fig.-1]. Being its location in theloop of Eastern Himalayas, it is more humid and has differenthydrological characteristics than the other parts of North-East
Region of the Country. The length of the main river isrecorded 145 km with an average slope of 5-15% with theperimeter of 264 km. Topographically, hill slopes are steep
covering a area of about 61.54% with its narrow flat valleys
located in the upper parts of the river catchment. Such
topography helps in accelerating the saturation processes andfast flow while the lower part is gentle plain accommodating
about a quarter part of the catchment [27.01%] with sedimentdeposition [Fig.-2].

Fig.-1: Location of the Dikrong river catchment867

Abbreviations: Slope in percent; 1= Very Gentle (2- 4), 2=
Gentle (4-10), 3= Moderate (10-20), 4= Moderately Steep(20-35), 5= Steep (35-60), 6= Very Steep (60-100), 7= Most
Steep (100-175), 8= Extremely Steep (above 175).
Fig-2: Slope Variations in the Dikrong River Catchment
Land use/ land cover pattern of Dikrong river catchment isdominated by forest [75% areas are under dense and open
forests] in the upper parts and agricultural and abandoned
land [12% of total area] in the lower parts of the catchment. In
such land surface conditions, the runoff has peculiar trend todevelop erratic floods in the lower parts of river catchment.
Soils are fine loamy and coarse silty which have high factionof sand helping in retaining more water to regulate runoff. As
a result, runoff has also been recorded in the dry weather of
winter seasons in spite of less rainfall and moderate PET
conditions [Fig.-3].

Fig.-3: Normal water balance for Dikrong river catchment
2. METHODS AND MATERIALS
Of course, there are various methods of assessing thereference evapotranspiration [ET 0 ] mainly classical method
forwarded by Blaney-Criddle establishing relation of day
time duration with temperature, Hamon method that is basedon water vapor density in the environment, Hargreaves andSamani on radiation balance and temperature variation, andPapadakis method which is saturation deficit based [12], [13],[14]. Such methods are largely dependent on fieldobservations used for different environmental conditions. The
PET calculation by T-M procedure which is first introducedby Thornthwaite [15] and, later on, Thornthwaite and Mather[16] with its notebook procedure has been used by many
geographers [17], meteorologists and climatologists [18], andfound its suitability especially for the monsoon basedhydrologic regime. In this method, temperature- based heatindex, i.e., i= [T0c/5]1.514 with a location specific correction
factor of unadjusted PET is used for the assessment of soil
moisture and runoff variability over time. In order to compare
daily runoff with other parameters of water balance equation[P= RO+PET+∆ST, where RO= Runoff, P= Precipitation,
PET= Potential Evapotranspiration and ∆ST= Changes in soil
moisture], the T-M procedure is used. In order to make data
standard, daily statistics of temperature and rainfall for threeconsecutive years [2004-2006], its average figures were used
for Itanagar station [that is centrally located in the study area].
The concerned statistics were collected from the Rural Works
Department, Government of Arunachal Pradesh, Itanagar.
While daily discharge statistics of the same period of time
were collected from the Water Resource Department,
Government of Assam, Guwahati for Sisapather gauge
station [that is at the mouth of river catchment]. Calculationsof daily soil moisture availability and runoff were made to set
the water holding capacity of 200 mm as specified by NBSS
& LUP, Jorhat [19] with adopting standard procedure ofwater retention capacity analysis.
3. RESULTS AND DISCUSSION
Of course, rainfall-runoff relationship always becomes
positive with its strong degree though with Curve Number[CN] method was adopted by USDA [20] to evaluate theeffect of land cover/land use and soil types on runoff.
Similarly, this relationship is significantly strong [r= 0.996]
with its regression gradient of slightly lesser than unity [1:1
relationship shows CN= 100 in the situation that there is noeffect of land use and soil on runoff generation in the rivercatchment] as b= 0.925 [Fig.-4]. The rainfall-runoff ratio istherefore calculated 0.5624 for Dikrong river catchment. Itmeans 56.24 % rainfall becomes runoff in this catchment andshows the effect of surface condition [slope, soil and land use]on runoff processes. Runoff regression equation showsinterestingly that daily rainfall below 1.24 mm does notproduce any runoff in this area as [a= -1.242] that is indicativeof the origin of runoff. Such runoff production level in theriver catchment clearly shows that there is production of
runoff during winter season varying from 4.80mm per day in
November to 4.24mm in January [Fig.-5]. Runoff of about
4.47mm daily was observed in these days due to sub-surface
flow, while its amount is influenced by soil and land use
factors. Concentration period of two days to reach the runoff
at mouth is also factor which shifts the runoff trend.

868

Fig.-4: The relationship between daily rainfall and daily RO

Fig.-5: Daily trends rainfall and observed RO
Rainfall does not correspond much to PET because PET ismore associated with temperature variation. The relationshipbetween them is too week [r= 0.475] to draw someconclusions. However, the degree of fluctuation of rainfall isobserved much higher [CV=124.78%] than PET [CV=47.9%] [Table-1].
Table-1: Daily Mean, Standard Deviation and Coefficient of
Variation of Hydrologic Parameters
Coefficientof
Variation
DailyStandard
Parameters
Mean
Deviation(%)
Mean daily
Temperature (T0 C)
22.57
4.19
18.58
Rainfall (mm)
14.44
18.02
124.78Heat Index (unit less)
9.92
2.72
27.38
PET (mm)
3.12
1.49
47.96
AE (mm)
2.33
1.98
84.85
Runoff (mm)
8.12
16.75
138.1
Runoff Ratio (%)
56.24-

In the event of increasing rainfall-intensity with wet
antecedent moisture conditions, the rainfall-runoff ratio
becomes higher that produces more discharge in the month oflate June to August. There are fast changes in soil moisture
conditions during the dry winter season when rainfall is
occasional that infiltrates its total amount and fluctuates soil
moisture availability. The summer does not have muchchange in antecedent soil moisture conditions. They are fully
saturated and generate more surface and subsurface runoff.
A total runoff is predicted 4,426mm annually with its
rainfall-runoff ratio of .8398 that deviates 27.74 % fromobserved runoff. Comparing observed with predicted runoff
over time, it is obvious to highlight that at the time of highrainfall intensity during the wet monsoon season, T-M
Procedure made over estimation of runoff trend and it wasunder-estimated during the dry winters. On the whole, theannual prediction of RO is over estimated as it predicts higher
rainfall-runoff ratio. However, the trend of observed runoff is
found much smoother than predicted one [Fig.-6] as the mainfeatures of such trends are given below.
i. Increasing rainfall increases fluctuation in runoff. For
example, the month of July and August [peakmonsoon season] has its very high fluctuation.
ii. On account of over estimation of runoff by T-M method,the error term in distribution becomes too high that
reduces the significance level with low degree ofdeterminant [R2=0.306] [Fig.-7].
iii. There is fast fluctuation in the soil moisture availability
during winter season when there is an occasional
rain. But on the other hand, PET trend is quitesmooth, increases gradually during pre-monsoon
months [April-May], becomes flat during monsoon
period and again fall down during the post-monsoon
period. PET follows the temperature variation rather
than rainfall trend [Fig.-8].
iv. The ratio of rainfall-runoff recorded very high during themonsoon period because of the fully saturated soil
and its wet antecedent soil moisture conditions.

Fig.-6: Daily trends of observed and predicted runoff869

trend of daily RO. Daily RO is not so smooth in this region as
it happens in the sub-tropical region of the world. Dry winters
have one-third discharge conditions in the river catchments
when rainfall was recorded very low. On the other hand, theamount of RO is concentrated during the monsoon season
when more that 80% rain is precipitated. There are somespecific findings inferred from the present analysis:a. No doubt, T-M procedure is simple to use withavailable parametric statistics of meteorologicalphenomena. However, it fails to predict correctly the
PET as well as RO parameters for this region of highhumidity.
b. T-M procedure estimate more fluctuation in daily
predicted RO phenomena than the observed one.
c. It over estimate the RO at high rainfall intensity.

Fig.-7: The relationship between daily observed
and predicted RO

Fig.-8: Trends of daily A. Changes in Soil Moisture, B.
Potential Evapotranspiration
4. CONCLUSION
In the specific high humid conditions of landforms in thefoot-hills of Arunachal Himalayas, where temperature varies
from 100-350C during the hydrologic regime of monsoon
climate, there are some general conclusions regarding the

5. REFERENCES[1] Tadesse N, “Surface Water Potential of the Hantebel
Basin, Tigray, Northern Ethiopia,” Agri. Engi. Int. CIGR
Ejournal, Vol. 8, 2006, pp. 1-31.
[2] Strzepek KM, Yates DN and Quosy DENE,“Vulnerability assessment of water resources in Egypt toclimatic changes in the Nile Basin,” Climate Research,
Vol. 6, 1996, pp. 89-95.
[3] Singh S and Syiemlieh HJ, “Prediction of runoff in the
Um-U-Lah catchment of extremely humid area ofCherrapunjii, India,” Geographia Polonica, Vol. 82 No.
2, 2011, pp. 127-139.
[4] Mishra SK, SArkar R, Dutta S and Panigrahy S, “A
physically hydrological model for paddy agriculture
dominated hill watershed in tropical region,” Journal of
Hydrology, Vol. 357, 2008. Pp. 389-404.
[5] Meshram DT, Mittal HK, Gorantiwar SD and Purohit
RC, “Reference crop evapotranspiration of western partof Maharashtra, India,” Indian J. Soil Cons., Vol. 38 No.
2, 2010, pp. 80-85.
[6] Lenka NK, Panda RK, Sudhishri S and Dass A,“Comparison of some empirical methods for estimatingof ET 0 in high lands of Eastern Ghats of Orissa,” Indian
J. Soil Cons., Vol. 37 No. 1, 2009, pp. 22-25.
[7] Blaney HF and Criddle WD, “Evaporation from freewater surface at high altitudes,” Trans. ASCE., Vol. 123,
1962, pp. 243-265.
[8] Hamon WR, “Estimating potential evapotranspiration,”
Journal of Hydraulics Division, Proceedings of theAmerican Society of Civil Engineers Vol. 871, 1961, pp.
107–120.
[9] Hargreaves GH, and Samani ZA, “Reference crop
evapotranspiration from temperature,” Trans. ASAE,
Vol 1 No. 2, 1985, pp. 96-99.
[10] Papadakis J, “Climate of the World and theirPotentialities,” A. V. Cordoba 4564, Buenos Aires,
Argentina, 1975, pp. 5-6.
[11] Beven K, “A sensitivity analysis of Penman-Monteithevapotranspiration estimates,” Journal of Hydrology,
Vol. 44, 1979, pp. 169-190.
[12] Xu, CY and Singh VP, (2001): “Evaluation andgeneralization of temperature-based methods for870

calculating evaporation,” Hydrol. Process, Vol. 15, 2001,
pp. 305–319.
[13] Loukas A, Vasiliades L and Mpastrogiannis N, “ Water
balance model for the evaluation of surface water
resources, in Proc. 8th Int. Conf. on Environmental andTechnology, 2003, pp. 492-499.
[14] Bautista F, Bautista D and Carranza CD, “Calibration of
the equations of Hargreaves and Thornthwaite toestimate the potential evapotranspiration in semi-aridand subhumid tropical climates for regional
applications,” Atmósfera, Vol. 22 No. 4, 2009, pp.
331-348.
[15] Thornthwaite CW, “An approach towards a rationalclassification of Climate,” Geogr. Rev., Vol. 38 No. 1,
1948, pp. 55-94.
[16] Thornthwaite CW and Mather JR, “Instructions andTables for Computing Potential Evapotranspiration andWater Balance,” Publications in Climatology, Vol. X No.
3, Drexel Institute of Technology, Centertown, New
Jersey, 1957, pp. 185-204.
[17] Soja R, Starkel L and Singh S, “Water Circulation and
Availability of Water,” Land Degradation and
Ecorestoration in the Extremely Wet Monsoon
Environment of Cherrapunji Area, India, 1st ed., Singh S,
Starkel L and Syiemlieh HJ Eds. Bookwell, New Delhi,
2010, pp. 30-50.
[18] Subrahmanyam, VP, “Water balance approach to thestudy of aridity and droughts with special reference to
India,” in Proc. of the Hamburg Workshop, 1983, pub.
No. 148, pp. 125-135.
[19] NBSS and LUP, “The Soils of Arunachal Pradesh,” at
R.F. 1: 250,000, prepared for Government of Arunachal
Pradesh, Itanagar by the National Bureau of Soil Survey
and Land Use Mapping, Regional Centre, Jorhat, 1993.
[20] USDA- Soil Conservation Service, Hydrology: National
Engineering Handbook, Washington DC, 1972, Section
4.

871

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Natural Radiation level in Groundwater from Katsina, a Semi Arid Region of
Northern Nigeria
B.G. Muhammad1 and M.S. Jaafar2
1. Department of Physics, Umaru Musa Yar’adua University, Katsina.
2. School of Physics, Universiti Sains, Malaysia.
e-mail: bashirgide@umyu.edu.ng
ABSTRACTThe World Health Organization’s radiological water quality protocol was observed to screen samples of groundwaterfrom Katsina, a semi-arid region part of northern Nigeria, for radiological safety. Initially, the gross alpha and grossbeta activity concentrations as measured by an eight channel gas flow proportional counting system indicated someparts of the study area to have elevated alpha and beta activity concentrations when compared with the ICRP (1991)
Maximum Acceptable Values (MAVs) of 500 Bqm-3 for alpha; 1,000 Bqm-3 for beta activity concentrations. This
prompts the need for further screening. In this respect, gamma spectroscopy was employed to evaluate the naturalradioactive materials responsible for the elevated alpha and beta activity concentration values. Subsequently,
UNSCEAR dose conversion factors (DCF) were used to estimate the committed effective dose so as to ascertain thehealth risk level of using the water for drinking purposes.
Keywords: natural radioactivity, activity concentration, committed effective dose.
1. INTRODUCTION
NORM, an acronym for naturally occurring radioactive
materials, refers to the long-lived radioactive elements suchas uranium, thorium and potassium and their decay products,
such as radium and radon that existed since the formation of
the earth. These elements have always been present in theearth's crust and subsequently, in all environmental media(soil, air and water). Their presence makes them inevitably
ingested into humans through food and drinking water [1-3].
Evidences from both human and animal studies also showsthat radiation exposure at low to moderate doses such as thosefrom drinking water may increase the long-term incidence ofcancer [4-6]. Due to the possible adverse effect fromingestion of radionuclides, through drinking water, a standard
need to be set in order to curb the unwanted radiation
exposure to human. Many countries adopt the WHO
recommendation of reference dose level [RDL] of committed
effective dose of 0.1 mSv from one year's consumption ofdrinking water[7]. The RDL of 0.1 mSv is equal to 10% of thedose limit for members of the public, recommended by the
ICRP (1991) and the International Basic Safety Standards(IAEA, 1996) and are acceptable to most WHO Member
States [8].
In arid and semi-arid regions, such as the study area,
evapo-transpiration rates are much high, recharge is less, flow
paths longer and residence times much greater and hencemuch higher levels of natural mineralization. Thus the major
ion contents and TDS are often high. In some desert regions,even if groundwater can be found it may be so salty(extremely high TDS) as to be undrinkable, and the difficultyof meeting even the most basic domestic requirements canhave serious impacts on health and livelihood. This prompts theneed to always screen domestic water supply from groundwater

sources that are often utilized in arid regions. The WHOradiological water quality protocol was followed to screen
samples of groundwater from Katsina, a semi-arid region ofnorthern Nigeria, for radiological safety. The result shows
elevated gross alpha and gross beta values exceeding the
WHO maximum contaminants levels of 500 and 1000 Bqm-3
respectively. The results of this segment of the work was
earlier reported[9]. In the present work, gamma spectroscopy
was employed to determine the natural radioactive materials
responsible for the elevated alpha and beta activity
concentration values. Subsequently, UNSCEAR doseconversion factors (DCF) were used to estimate thecommitted effective dose so as to ascertain the health risk
level of using the water for drinking purposes.
Materials and Methods
Sampling
After using a GPS to measure the geographical coordinates ofthe sampling sites, clean polythene containers were used tocollect groundwater from boreholes of the selected sites. The
sites were chosen for their proximity to the neighboring Niger
Republic that is known to have natural uranium deposits as
depicted in the IAEA world’s commercial uranium deposits
map. The samples were then immediately brought to thelaboratory for analysis. Guidelines provided by internationalorganization for standardization on water quality sampling
were strictly adhered to[10]. To prepare water samples for
gamma spectroscopy, 500 ml of each sample was acidified
with 10 ml of concentrated HNO 3 (analytical grade) [11] andthen sealed in the polythene counting sample container andkept for a period of at least four weeks prior to counting.
Gamma spectroscopic system
Gamma spectroscopic measurements were carried out using ahigh efficiency 3’’x 3’’ Thallium doped, sodium iodide[NaI(Tl)] scintillation detector (Ortec, USA), with serial872

number 101909H connected to a 3’’ (B76D01W N 2034257)
PMT photomultiplier tube. The set-up was coupled to a multichannel analyser (MCA). The detector was enclosed in a 10
cm lead shield to reduce background radiation in the counting
environment. The set-up is illustrated in figure 1 below. The
system has relative efficiency and resolution of about 50%
and 7.2% respectively at energy of 0.662 MeV (137Cs). These
are considered adequate to distinguish the gamma rayenergies of interest in this work. The spectral analysis andpeak fitting were performed using Fitzpeak NaI(Tl) (lowresolution) gamma calibration and analysis softwaredeveloped by JF Scientific UK. This software is specially
optimized to be used for low resolution gamma detector
NaI(Tl). The energy calibration was carried out using IAEA
certified standards of known energies such as 60Co, for E γ =1332.5 and 1173.2 keV, and 137Cs, for E γ = 661.6 keV. The
efficiency calibration was also performed by using IAEA
certified 152Eu standard[12] for E γ = 121.8, 224.7, 344.3,
778.9, 964.0, 1085.8, 1112.0 and 1408.0 keV. A counting
time of 86400 s was allowed for both background and sample
due to the low emission nature of the samples. The activity
concentrations of the radionuclides in the samples were
determined from their respective gamma-ray lines or
gamma-ray lines emitted from their decay products. The
gamma line of 1460.8 keV was used directly todetermine 40K. The weighted mean activity concentrations
from gamma lines of 911.1 keV (228Ac) and 583.1 keV(212Pb) were used to determine 232Th. The gamma lines of
351.9 keV (214Pb) and 609.3 keV (214Bi) were used todetermine 238U and 186.1 keV (226Ra) gives the activity
concentration of 226Ra as a natural source.
Lead Shield

PC withFitzpeak(NaI) low
resolution
softwareSampleNaI(Tl)
PMT

HV
Power
SupplyInput

AmplifierOutputMCA

Fig. 1 The radio-elemental gamma spectroscopy set-up
Committed effective dose
Estimation of committed effective dose E d (Svy-1) to anindividual due to the consumption of the naturalradionuclides present in the drinking water was done using
equation 1 [1]. The WHO A i values of 230, 330 and 730 litresfor infants (1-2 years), children (˃7-12 years) and adults (˃17years) respectively were used [7, 13].
E d = Ac* Ai * C fwhere A c is the radionuclide activity concentration (Bql-1), A i
is the annual intake of water (ly-1) and C f is the doseconversion factor for the radionuclide (SvBq-1). The doseconversion coefficients provided by UNSCEAR 2000
report[14] were also used for the dose estimation. The totalcommitted effective dose D (Sv y-1) to an individual wasestablished by summing up the contributions from allradionuclides present in the water samples as given byequation 2.
D = ∑ Ac* Ai * C f(2)

Results and Discussion
The activity concentration of natural radionuclides in the
groundwater from the study area is presented in table 1. Table
2 presents the estimated total committed effective dose
resulting as a consequence of direct ingestion of dissolved
radionuclides in the groundwater. Table 3 compares theresults of the present study with those from other parts of theworld.
The world annual natural radioactivity level, including
external exposure, consumption of food and water containing
natural radionuclides, and inhalation of radon with its
daughter products, amounts on average to 2.4 mSv, with arange from 1 to 10 mSv[19]. The United Nations ScientificCommittee on Effects of Atomic Radiation (UNSCEAR)
report also indicated that the average worldwide exposure tonatural sources in foods and drinking water is 0.29 mSv y-1(about 0.17 mSv y-1 from 40K and about 0.12 mSvy-1 from
Uranium and Thorium)[19]. For prolonged exposure
situations in case of all environmental radioactive sources,
including natural and human-made sources, the International
Commission on Radiological Protection (ICRP) recently
recommended a generic intervention level of annual dose of
10 mSv as the level below which intervention is unjustified,taking into account radiological, economic and social
factors[20]. The world health organization in its guidelines fordrinking water quality, adopted the International commission
on radiological protections’ recommendation of reference doselevel (RDL) of the committed effective dose, to be equal to
0.1mSv from 1 year’s consumption of drinking-water (fromthe possible total radioactive contamination of the annual
drinking-water consumption).
Comparing the results of this study with other sources fromother parts of the world, the mean measured activity
concentration values of 0.43, 0.52 and 1.02 Bq l-1
for 232Th, 40K and 226Ra respectively are well below thereported concentrations of 12, 12 and 29 Bq l-1 of (untreated)lake water from Biseni, Nigeria[17]. In water used for
drinking and other domestic purposes, the mean measured
activity concentration values on the other hand, are above theearlier reported values for bottled water from Pakistan as wellas the tap water from Egypt, but the values (226Ra and 40K)
from this area are still far below the earlier reported values
for well drinking water from Akure Nigeria, where high levelof radium was discovered. Consequently, the total committed

(1)
873

Table 1 Activity concentration of natural radionuclides in the groundwater sources
S/N
1.
2.
3.
4.
5.
6.
7.
8.
9.
10.
11.
12.
13.
14.
15.
16.
17.
18.
19.
20.
21.
22.
23.
24.
25.
26.
27.
28.
29.
30.
31.
32.
33.
34.
35.
36.
37.
38.
39.
40.

Location
Location
Geographicalcoordinates
Dutsin Safe Low-cost
N13o 00′ 20.1″
Dutsin Safe extension
N13o 00′ 11.7″
Fed. Medical Centre
N12o 59′ 06.4″
Sokoto Rima
N12o 58′ 41.5″
AbbatoirN12o 58′ 24.3″
Central Market
N12o 58′ 16.3″
Sabuwar Unguwa
N12o 58′ 11.0″
Bread
N12o 58′ 23.1″
Garama Quarters
N12o 59′ 07.9″
Sabon Layi
N12o 59′ 13.0″
Emir’s Palace
N12o 59′ 42.5″
Farin Yaro
N12o 59′ 48.3″
Unguwar Yari
N12o 59′ 28.9″‘Yar adua
N12o 59′ 38.7″
Tudun Wada
N12o 59′ 54.4″
School of Nursing
N13o 00′ 04.2″
Filin Samji
N13o 00′ 22.4″
Dutsin Amare
N12o 59′ 30.2″
Kofar Marusa
N12o 59′ 19.1″
Shararrar Pipe
N12o 58′ 38.1″
Rafin Dadi
N12o 58′ 57.0″
Dan Marna
N12o 59′ 06.2″
LayoutN12o 58′ 56.5″
W.T.C.
N12o 58′ 34.4″
New StadiumN12o 58′ 17.0″
C.B.N.
N12o 59′ 24.1″
High Court
N12o 59′ 12.5″
Motel
N13o 00′ 01.8″
G.R.A.
N12o 59′ 59.8″
Cutoms Barracks
N13o 00′ 23.8″
Goruba Housing Est.
N13o 00′ 06.9″
Nitel
N12o 59′ 47.3″
Kofar Durbi
N12o 59′ 42.2″
Social Development
N12o 59′ 55.2″
Area Council
N12o 59′ 42.7″
Unguwar Liman
N12o 59′ 56.8″
Kofar Kaura
N12o 59′ 56.8″
Canada Quarters
N12o 58′ 16.1″
Aishatu Memorial
N12o 58′ 07.8″
Danfodio InstituteN12o 58′ 40.2″

Activity Concentration (Bq l-1)
232
40
Th
K

238

U

E7o 35′ 10.1″
E7o 35′ 00.9″
E7o 34′ 38.8″
E7o 35′ 37.3″
E7o 35′ 27.7″
E7o 36′ 06.0″
E7o 36′ 32.7″
E7o 36′ 50.4″
E7o 36′ 19.9″
E7o 36′ 19.1″
E7o 36′ 01.0″
E7o 35′ 30.5″
E7o 35′ 36.0″
E7o 36′ 18.7″
E7o 36′ 26.4″
E7o 36′ 16.3″
E7o 36′ 35.4″
E7o 36′ 39.6″
E7o 36′ 33.2″
E7o 36′ 23.5″
E7o 35′ 40.5″
E7o 36′ 05.6″
E7o 36′ 56.5″
E7o 37′ 50.1″
E7o 38′ 34.5″
E7o 37′ 47.2″
E7o 38′ 22.5″
E7o 37′ 53.5″
E7o 38′ 28.6″
E7o 37′ 19.1″
E7o 37′ 16.8″
E7o 37′ 08.6″
E7o 36′ 57.7″
E7o 36′ 51.7″
E7o 36′ 33.0″
E7o 35′ 53.5″
E7o 36′ 17.3″
E7o 37′ 07.3″
E7o 37′ 24.2″
E7o 37′ 16.5″

0.48±0.035
0.49±0.028
0.22±0.011
0.29±0.021
0.22±0.012
0.34±0.021
0.31±0.017
0.78±0.076
0.81±0.065
0.39±0.022
0.50±0.031
0.44±0.062
0.04±0.003
0.47±0.053
0.43±0.033
0.43±0.043
0.46±0.036
0.42±0.021
0.04±0.032
0.49±0.033
0.46±0.023
0.44±0.022
0.46±0.034
0.06±0.027
0.39±0.022
0.39±0.034
0.39±0.055
0.39±0.042
0.41±0.044
0.85±0.058
0.18±0.014
0.85±0.043
0.77±0.062
0.85±0.066
0.73±0.056
0.60±0.034
0.77±0.055
0.35±0.034
0.46±0.035
0.42±0.051

0.39±0.021
0.54±0.033
1.30±0.023
0.17±0.011
0.04±0.003
0.23±0.012
0.27±0.023
0.64±0.043
0.66±0.017
0.15±0.014
0.49±0.051
1.20±0.087
1.11±0.096
0.24±0.019
0.12±0.023
0.14±0.012
0.33±0.023
0.21±0.018
0.02±0.003
0.30±0.016
0.22±0.017
0.16±0.021
0.20±0.022
0.05±0.002
0.23±0.011
0.26±0.021
0.27±0.015
0.20±0.009
0.25±0.021
0.07±0.004
0.20±0.021
0.97±0.055
0.89±0.037
0.89±0.054
0.83±0.076
0.67±0.044
0.70±0.034
0.86±0.066
0.33±0.044
0.21±0.031

226

Ra

0.79±0.042
0.78±0.067
0.91±0.072
1.16±0.091
0.91±0.082
1.32±0.088
1.20±0.076
0.39±0.031
4.07±0.089
0.64±0.067
0.89±0.073
0.78±0.055
0.71±0.076
0.71±0.081
0.70±0.073
0.68±0.056
0.78±0.055
0.73±0.047
0.60±0.044
0.84±0.067
0.73±0.056
0.72±0.066
0.77±0.043
0.73±0.052
0.71±0.055
0.65±0.061
0.62±0.034
0.64±0.033
0.72±0.045
1.32±0.086
3.00±0.098
1.56±0.077
1.32±0.087
1.32±0.078
1.32±0.088
0.95±0.065
1.32±0.073
1.19±0.077
0.78±0.045
0.73±0.052

Table 2 Estimated age dependent total committed effective dose (mSv y-1) in groundwater
due to natural radionuclides in the area
S/N

Location

Geographical coordinates

1.
2.
3.
4.
5.
6.
7.
8.
9.
10.

Dutsin Safe Lowcost
Dutsin Safe extension
Fed. Medical Centre
Sokoto Rima
AbbatoirCentral Market
Sabuwar Unguwa
Nakowa Bread
Garama Quarters
Sabon Layi

N13o 00′ 20.1″
N13o 00′ 11.7″
N12o 59′ 06.4″
N12o 58′ 41.5″
N12o 58′ 24.3″
N12o 58′ 16.3″
N12o 58′ 11.0″
N12o 58′ 23.1″
N12o 59′ 07.9″
N12o 59′ 13.0″

E7o 35′ 10.1″
E7o 35′ 00.9″
E7o 34′ 38.8″
E7o 35′ 37.3″
E7o 35′ 27.7″
E7o 36′ 06.0″
E7o 36′ 32.7″
E7o 36′ 50.4″
E7o 36′ 19.9″
E7o 36′ 19.1″

874

Infants(1-2 y)
0.14
0.16
0.42
0.39
0.30
0.46
0.43
0.46
0.51
0.12

Children(2-16 y)
0.14
0.15
0.43
0.41
0.31
0.49
0.46
0.49
0.52
0.10

Adults(˃16 y)
0.15
0.18
0.47
0.35
0.26
0.42
0.39
0.46
0.49
0.10

0.29±0.012
0.28±0.021
1.09±0.073
1.39±0.077
1.09±0.081
1.68±0.093
1.54±0.082
1.48±0.076
1.55±0.088
0.22±0.033
0.32±0.023
0.28±0.022
0.25±0.012
0.25±0.018
0.25±0.021
0.25±0.014
0.28±0.022
0.27±0.017
0.22±0.011
0.31±0.021
0.27±0.015
0.27±0.023
0.28±0.023
0.27±0.011
0.25±0.031
0.24±0.014
0.22±0.017
0.24±0.021
0.27±0.033
0.48±0.034
1.05±0.088
0.53±0.045
0.46±0.033
0.45±0.034
0.45±0.025
0.35±0.026
0.46±0.044
0.42±0.031
0.28±0.013
0.27±0.016

11.
12.
13.
14.
15.
16.
17.
18.
19.
20.
21.
22.
23.
24.
25.
26.
27.
28.
29.
30.
31.
32.
33.
34.
35.
36.
37.
38.
39.
40.

Emir’s Palace
Farin Yaro
Unguwar Yari‘Yar adua
Tudun Wada
School of Nursing
Filin Samji
Dutsin Amare
Kofar Marusa
Shararrar Pipe
Rafin Dadi
Dan Marna
LayoutW.T.C.
New Stadium
C.B.N.
High Court
Motel
G.R.A.
Cutoms Barracks
Goruba Housing Est.
Nitel
Kofar Durbi
Social Development
Area Council
Unguwar Liman
Kofar Kaura
Canada Quarters
Aishatu Memorial
Danfodio InstituteN12o 59′ 42.5″
N12o 59′ 48.3″
N12o 59′ 28.9″
N12o 59′ 38.7″
N12o 59′ 54.4″
N13o 00′ 04.2″
N13o 00′ 22.4″
N12o 59′ 30.2″
N12o 59′ 19.1″
N12o 58′ 38.1″
N12o 58′ 57.0″
N12o 59′ 06.2″
N12o 58′ 56.5″
N12o 58′ 34.4″
N12o 58′ 17.0″
N12o 59′ 24.1″
N12o 59′ 12.5″
N13o 00′ 01.8″
N12o 59′ 59.8″
N13o 00′ 23.8″
N13o 00′ 06.9″
N12o 59′ 47.3″
N12o 59′ 42.2″
N12o 59′ 55.2″
N12o 59′ 42.7″
N12o 59′ 56.8″
N12o 59′ 56.8″
N12o 58′ 16.1″
N12o 58′ 07.8″
N12o 58′ 40.2″

0.18
0.21
0.18
0.10
0.10
0.10
0.12
0.11
0.08
0.15
0.12
0.11
0.12
0.07
0.10
0.10
0.10
0.09
0.10
0.15
0.29
0.26
0.23
0.24
0.22
0.18
0.22
0.26
0.12
0.11

E7o 36′ 01.0″
E7o 35′ 30.5″
E7o 35′ 36.0″
E7o 36′ 18.7″
E7o 36′ 26.4″
E7o 36′ 16.3″
E7o 36′ 35.4″
E7o 36′ 39.6″
E7o 36′ 33.2″
E7o 36′ 23.5″
E7o 35′ 40.5″
E7o 36′ 05.6″
E7o 36′ 56.5″
E7o 37′ 50.1″
E7o 38′ 34.5″
E7o 37′ 47.2″
E7o 38′ 22.5″
E7o 37′ 53.5″
E7o 38′ 28.6″
E7o 37′ 19.1″
E7o 37′ 16.8″
E7o 37′ 08.6″
E7o 36′ 57.7″
E7o 36′ 51.7″
E7o 36′ 33.0″
E7o 35′ 53.5″
E7o 36′ 17.3″
E7o 37′ 07.3″
E7o 37′ 24.2″
E7o 37′ 16.5″

0.16
0.20
0.18
0.10
0.10
0.10
0.12
0.11
0.07
0.14
0.11
0.11
0.11
0.08
0.10
0.10
0.10
0.09
0.11
0.16
0.31
0.26
0.23
0.23
0.22
0.18
0.21
0.29
0.12
0.11

0.19
0.28
0.24
0.11
0.10
0.10
0.14
0.11
0.06
0.15
0.12
0.11
0.12
0.07
0.11
0.11
0.11
0.10
0.11
0.14
0.27
0.31
0.28
0.28
0.26
0.21
0.25
0.25
0.14
0.11

Table 3 Natural radioactivity in water from other parts of the world
Area
Egypt
Islamabad, Pakistan
Biseni, Nigeria
Akure, Nigeria
Chittagong, Bangladesh
Katsina, Nigeria

232

Type of Source
Tap water
Bottled waterLake water
Well water
Well and tap
Groundwater

Th
0.008
0.0113
12
NR
0.19
0.43

NR- not reported

effective dose due to one year’s direct ingestion of thegroundwater from the study area was estimated to vary from0.07 to 0.46, 0.08 to 0.49 and 0.06 to 0.49 for infants, children
and adults respectively. This indicates that the value in mostof the groundwater sources seems to exceed the reference
dose values of 0.1 mSv y-1 as recommended by the WHO.
However, the total committed effective dose estimated in all
the samples considered is below 1 mSv y-1, which is the 10%of the level above which intervention is justified taking into
account radiological, economic and social factors.
CONCLUSION
The results of this study provides a baseline data of thenatural radioactivity levels in the groundwater resources
abundantly utilised for drinking and other domestic purposesby inhabitants of Katsina, a semi arid region in northern
Nigeria. Comparatively, the moderate natural activity

Mean specific activity (Bq l-1)
226
40
Ra
K
Reference
0.0157
0.152
[15]
0.052
0.0409
[16]
12
29
[17]
7.15
13.54
[18]
0.043
4.16
[1]
0.52
1.02
This work

concentration values in groundwater from arid and semi arid
regions such as the study area are attributable to the nature of
the water sources, that is, less recharge and high residence
time resulting in more dissolved solids accumulation in thewater. Despite the fact that the concentrations of naturalradioisotopes and subsequent committed effective doses inmany samples were substantially higher than those in otherpopulations, most still remained low when compared withgeneric intervention level. The magnitude of risk expected inour study is therefore small. Radionuclides screening as wellas the use of best available technologies (BAT) such as ion
exchange, reverse osmosis or lime softening (asrecommended by USEPA) for areas where the naturalradionuclides concentrations results in committed effective
dose that exceeds RDL are recommended in order tosafeguard the members of the public from radiation exposure
above recommended limits.

875

REFERENCES
1.

2.
3.

4.
5.

6.
7.
8.
9.

10.
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16.

Alam, M.N.C., M.I. Kamal,M. Ghose,S.
Islam,M.N. and Anwaruddin,M., Radiologicalassessment of drinking water of the chittagong
region of Bangladesh Radiation Protection
Dosimetry, 1999. 82(3): p. 207-14.
Dalma, N., et al., Gross alpha and beta activities intap waters in Eastern Black Sea region of Turkey.
Chemosphere, 2006. 62: p. 956-960.
Gruber, V., F.J. Maringer, and C. Landstetter, Radonand other natural radionuclides in drinking water inAustria: measurement and assessment. Appl Radiat
Isot, 2009. 67(5): p. 913-7.
Abo, M., et al., Cytogenetic effects of radon
inhalation. Radiation Measurements, 2008. 43: p.
1265 - 1269.
Collman, G.W., D.P. Loomis, and D.P. Sandler,Childhood cancer mortality and radon
concentration in drinking water in North Carolina.
Br J Cancer, 1991. 63(4): p. 626-9.
Amrani, D. and D.E. Cherouati, Health effects fromradon-222 in drinking water in Algiers. J Radiol
Prot, 1999. 19(3): p. 275-9.
WHO, Guidelines for Drinking-water Quality.
World Health Organisation Publiation, 1988. 1: p.
197-209.
Bonotto, D.M. and T.O. Bueno, The naturalradioactivity in Guarani aquifer groundwater,
Brazil. Appl Radiat Isot, 2008. 66(10): p. 1507-22.
Muhammad, B.G., M.S. Jaafar, and T.C. Akpa, A
survey of gross alpha and beta activity
concentrations in groundwater from Katsina area ofNorthern Nigeria. Radiat Prot Dosimetry, 2010.
141(2): p. 127-33.
ISO, Water quality - Sampling - Part 1: Guidance on
the design of sampling programmes and samplingtechniques British Standard, 2006. 1(5667): p. 1-31.
ISO, Water quality - Sampling - Part 6: Guidance on
sampling of rivers and streamsBritish Standard,
2005. 6(5667): p. 1-16.
ISO, Measurement of radioactivity in theenvironment, in Measurement of gammaemitting
radionuclides. 2007, ISO. p. 30.
Ajayi, O.S. and J. Achuka, Radioactivity in drilled
and dug well drinking water of Ogun stateSouthwestern Nigeria and consequent dose
estimates. Radiat Prot Dosimetry, 2009. 135(1): p.
54-63.
UNSCEAR, Dose assessment methodologies, in
SOURCES AND EFFECTS OF IONIZINGRADIATION. 2000, UNSCEAR. p. 63.
Amin, R.M., F.A. Khalil, and M.A. El Fayoumi,
Natural radioactivity and chemical concentrationsin Egyptian groundwater. Environ Monit Assess,2011. 173(1-4): p. 29-35.
Fatima, I., et al., Measurement of naturalradioactivity in bottled drinking water in Pakistan

17.

18.

19.
20.

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and consequent dose estimates. Radiat Prot
Dosimetry, 2007. 123(2): p. 234-40.
Agbalagba, E.O. and R.A. Onoja, Evaluation ofnatural radioactivity in soil, sediment and water
samples of Niger Delta (Biseni) flood plain lakes,
Nigeria. J Environ Radioact, 2011. 102(7): p.
667-71.
Ajayi, O.S. and T.P. Owolabi, Determination ofnatural radioactivity in drinking water in private
dug wells in Akure, Southwestern Nigeria. Radiat
Prot Dosimetry, 2008. 128(4): p. 477-84.
UNSCEAR, Sources and effects of ionizing
radiation, UNSCEAR, Editor. 2000: New York.
Wrixon, A.D., New ICRP recommendations. Journalof Radiological Protection, 2008. 28(161): p.
161-168.

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Cost-Performance Studies of Abandoned Shell Husks (ASH) in Japan
Md. Toriqul Islam, Md. Zakaria Hossain, Masaaki Ishida, Md. Bellal Hossain and Md, Jamal Uddin
Graduate School of Bioresources, Mie University, Japan
Capital Development Authority (RAJUK), Bangladesh
ABSTRACTThis paper presents a comprehensive study on the cost analysis of abandoned shell husks (ASH) and it
performance for soil reinforcement applications. Along with field visit at the shell husk abandoned area,
data collected from the public city office and private company offices were analyzed. It was observed thatthe abandoned shell husks causing environmental pollution through spreading bad smell to the neighbor of
the abandoned area. Analyses of the data showed that the shells were not only abandoned without any
commercial value but huge amount of money was being spent for its disposal. In order to conserve natural
resources and to mitigate the pollution and environmental problems, composite reinforcement using
abandoned shell husks were prepared in the laboratory and soil-structure interaction tests were conducted forsoil reinforcement applications. Overall response of the shear properties of composite reinforcement
containing abandoned shell husk were demonstrated under varying surcharge such as different normal
loading conditions. It was observed that the use of shell husk in composite reinforcement improved the shear
performance significantly.
Keywords: Shell husk, Cost analysis, Environment pollution, Composite reinforcement, Slope protection

1. INTRODUCTION
A huge amount of abandoned shell husk is produced in Japanproduced as a byproduct of shell production every year casing
serious environmental problem through spreading bad smell
to the surroundings. Some abandoned shell husk thrown in
Tsu city, Mie Prefecture, Japan is shown in Fig.1. Along withenvironmental concern, a lot of money is being spent for the
disposal of these shell husks in Mie and all over Japan. With
the rapid increase in demand for balance between natural
phenomena and ecology in bio-environment, there is asignificant need for continuous development of newtechnologies that consume these abandoned shell husks [1,2].
This solves mainly two environmental and economic aspects,
such as, (1) solving the waste storage problem and (2)protection of limited natural resources of aggregates [3,4].
Both the economic and environmental benefits can beachieved by using the abandoned shell husk as recycled
aggregate in concrete or in cement-based composites. Some
researches on concrete or cementitious composites with otherrecycled aggregate produced from building destruction can befound in the technical literature [5-8]. Light weight concrete
masonry with recycled wood aggregate was studied by Stahl[9]. The latest researches on the assessment of the surface
resistance and permeation properties of recycled aggregatesconcrete is also found [10,11]. In spite of the volume ofavailable information in the technical literature, there is a noresearch work available on the use of abandoned shell husks
in cementitious composites for soil reinforcement
applications [12].

If abandoned shell husk properly applied for the surface
treatment of composite reinforcement, shear resistance
between the soil and the reinforcement may be significantly
improved. However, this needs to be substantiating by
evidence through laboratory experiment and if necessary by
field applications. In view of the above distinct advantages
for environmental conservation and protection of limited
natural resources as well as soil-reinforcement applications;the present research work is undertaken to fulfill this basic
need. It is expected that the outcomes of this research assist in
possible utilization of abandoned shell husk as aggregate for
producing cement-based composite reinforcements forprospective uses in the field such as; terrace lands, roads,
canals, pavements, embankments, ridge between paddy fields
and other agricultural and engineering structures in the formof earth slope protection. For effective utilization of
cementitious composites containing abandoned shell husks insoil and geotechnical works, it is necessary to study various
engineering properties such as soil-reinforcement interaction.
Shear tests are commonly performed to study the behavior
and shear resistance of cement based composite materials.
Shear tests (soil-structure interaction) between soil andcomposite reinforcement containing abandoned mactridae
shell husk under different surcharge such as normal loading
conditions were demonstrated. A comparison of the test
results with ordinary reinforcement is depicted.

877

Seashore of Tsu city

Abandoned shell husks
Fig.1 Abandoned mactridae shell husks in Tsu city, Mie Prefecture, Japan
2. DETAILS ABOUT ABANDONED SHELL HUSKThe abandoned shell husks were collected from the sea coastnear to Mie University, Tsu city, Mie Prefecture, Japan(Fig.1) are the Mactridae type shell husk. After collection, theshell husks were graded by performing sieve analysis. The
fineness modulus and the maximum size of the abandoned
shell husks were 4.35 and 4.76mm, respectively. The shellsize distribution curve can be found elsewhere []. According
to the Japanese Ministry of Forestry, Fisheries andAgriculture; Mactridae, also known as trough shells or duckclams, is a family of marine bivalve clams of the order
Veneroida. The amount of shell production and disposal costwere given in Table 1 based on city, prefecture and country
and depicted in the results and discussion section. Therefore,
there are not repeated in this section.
3. PREPARATION OF SPECIMENS
Ordinary Portland cement and river sand passing through aNo.8 (2.38mm) sieve, which has a fineness modulus of 2.33,
were used for casting. For all the specimens, the water tocement ratio and cement to sand ratio both were 0.5 by weight.
The requisite amount of sand and cement was dry-mixed in apan, and then the requisite quantity of water was added

gradually while the mix was continuously stirred.
Rectangular specimens were made in the wooden moulds
with open tops. The moulds were made in such a way that theside walls and the base were detachable so that the mould
could be easily separated from the specimens after its initial
setting (Fig.2). The contact surfaces of the mould to themortar were greased before casting the specimens to ease thedemolding process At first, thin layer (4 to 5 mm) of mortar
was spread on the base of the mold and then the mesh made ofsmall diameter (1mm) with c/c opening 10mm was placed on
the mortar. After that the second layer of mortar was spreadon the mesh covering by mortar again 4 to 5 mm. Therefore,
the final thickness of the composite reinforcement was madeof 10mm. The shell husks were spread on the mortar in such a
way so that each shell was 50% inside the mortar with 50%
above the mortar. The husks were spread randomly. Overallsurface area that was covered by the shell was 20%. A hole atthe center of the composite was made in order to drain porewater from the soil structure (Fig.3). This also provides
additional shear resistance.
The specimens were air-dried for 24 hours for initial setting
and then immersed in water for curing. The specimens were
removed from water after 28 days and were air-dried for 2days in room temperature of about 25oC and relative humidityof about 50%; then the tests were performed.

878

the lower box, the upper part was set on the panel, and thenthe Toyoura sand was rained in the upper box with 80cm
height. This provides nearly 90% compaction of the soil in theupper box. The tests were carried out in the way of pushingout the panel along with the lower box from the soil withconstant selected speed by means of screw jack underelectrically operated constant pressure. The shear force andthe displacements were measured by using the linear variable
differential transformers (LVDTs). All the tests were carriedout under surcharges of 40, 80, 120, 160 kPa.

Fig. 2 Mold for cast composite reinforcement

Fig. 4 Shear testing apparatus
5. RESULTS AND DISCUSSION

Fig.3 Completed composite reinforcement containing ASH
4. METHOD OF TESTING
The apparatus used in this study is shown in Fig. 4 which iscapable of performing both pullout and direct shear tests. For
convenience of the readers, the important components of thetesting equipment are numbered numerically starting from
top-left to right-down in the increasing way such as, thenumber from [1] to [8] where the number [1] is the normalload application plate for upper box, [2] is the shear stressmeasuring device, [3] is the upper box filled with soil, [4] is
the ASH panel, [5] is the lower box, [6] is the electricallyoperated shear jack, [7] is the displacement measuring dialgauge and [8] is the device taking normal load which acted on
the upper box.
TheASH panels were made to obtain rectangular pieces of
316 mm by 380 mm in size with 120-mm extended mesh. The
specified length of the pieces was selected in order tofacilitate clamping with the shear apparatus. The panels wereclamped in the box in such a way that the embedded length ofthe panel is 380 mm in the loading direction and 316 mm in
the transverse direction. After embedding the ASH panel on

The amount of ASH production and required cost for its
disposal in Japan are given in Table 1. As can be seen, theamount of the ASH is about 151000 tons per year according
to the Japanese Ministry of Forestry, Fisheries andAgriculture. Among these, about 9627 tons/year are producedin Mie prefecture and 715 tons/year are produced in Tsu city
along with 407 tons/year of Mactra Chinensis (Mactridaes)in Tsu city. According to the information of disposal site
office in Tsu city, the cost for disposing the ASH is about 18
yen per kilogram which means that nearly 12.8 million yen isrequired for disposing ASH in Tsu city (nearly 7.3 million
yen is required for Mactridaes) only. In addition to this,nearly 2 million US$ is required in Mie Prefecture and nearly
32 million US$ is required in Japan. These ASHs which are
not only thrown away without any commercial return and a
lot of money is being spent for its disposal but also causing
pollution and environmental problems.
Table 1 Amount of ASH and cost for its disposal in 2009
Place
Japan
Mie prefecture
Tsu City
Tsu City(mainly Mactra
Chinensis)879

Amount of shells
production (tons))
151,000
9,627
715

Cost required todispose shells (yen)2718000000
173286000
12870000

407

7326000

The shear behavior of ASH in terms of soil-ASH
interaction is depicted in Fig. 5 with the surcharges inabscissa and ultimate shear strength in ordinate.

4 Conclusions
The possible utilization of abandoned shell husks inconcrete construction is studied in this paper. Engineering
properties such as compressive strength, Young’s modulus,
tensile strength, unit weight, water absorption capacity andcoefficient of hydraulic conductivity results agree well with
the ACI results. In designing light weight concrete structuresuch as partition wall, slope surface protection, ridge between
paddy fields, roads and embankments slopes and sea-shore
protection structures where strength does not a great factor
but the water absorption properties are advantageous,concrete with shell aggregates may be used.

[10] Yamauchi, T. Sahara, H. and Kudo, K. On the Use of
Hotate Shells as Concrete Aggregates, Proceedings of the60th Technical Conference of Civil Engineering, 2005:
435-437 (In Japanese)[11] Frondistou-Yannas, S.T. Waste Concrete as aggregate
for New Concrete, ACI Journal, 1977; 74(8): 373-376.
[12] Hansen, T.C. and Boegh, E. Elasticity and Drying
Shrinkage of Recycled Aggregate Concrete, ACI Journal,
1985; 82(5): 648-652.
[1] Islam MR, “Conference proceedings,” in Proc. 2nd Int.
Conf. on GEOMAT, 2011, pp. 8-13.

References[1] Wang, I., Takamura, S., Nakamura, M. and Tsukinaga,
Y. A Study on the ratio and composition of the porous
concrete containing shells, Transactions of Cement andConcrete, 2003, 57:572-577 (In Japanese)[1] Delwar M. et al. Use of reclaimed asphalt pavements as
an aggregate in Portland cement concrete. ACI Material J,
1997; 94(3): 251-256.
[2] Kansai, Y. Demolition and reuse of concrete masonry,
Proceedings RILEM Symposium, Tokyo, Chapman and Hall,
1988, pp.774.
[3]
Hansen,T.C.
Recycledaggregatesandrecycled-aggregatesconcrete,state-of-the-art
reportdevelopments 1945-1985, RILEM Technical Committee
37-DRC, Material structures 1986, 19(111): 20-46.
[4] Rim, K.A. et al. Influence of the proportion of wood onthethermalandmechanicalperformancesofclay-cement-wood composites, cement and concrete
composites, 1999; (21): 269-276.
[5] Ajdukiewicz, A. et al. Influence of recycled aggregateson the mechanical properties of HS/HPC, cement andconcrete composites, 2002; (24): 269-279.
[6] Banthia, N. and Chan, C. Use of Recycled Aggregate inplain and Fiber Reinforced Shot Crete, ConcreteInternational: design and Construction, American Concrete
Institute, 2000; 22(6): 41-46.
[7] Sagoe-crentsil, K.K. et al. Performance of Concrete
made with Commercially Produced Coarse Recycled
Concrete Aggregate, Cement and Concrete Research, 2001;
31(5): 707-712.
[8] Hossain, M.Z. and Inoue, S. Finite Element Analysis ofThin Panels Reinforced with a Square Mesh, J. of
Ferrocement, 2002; 32(2): 109-125.
[9] Stahl [9]. D.C et al. Light weight concrete masonrywith recycled wood aggregate , J. of Materials in Civil
Engineering, 2002; 14 (2): 116-121.
880

Second International Conference on Geotechnique, Construction Materials and Environment,
Kuala Lumpur, Malaysia, Nov. 14-16, 2012, ISBN: 978-4-9905958-1-4 C3051

Cyclones Path and Severities: Effect on Flood Height and Damages of Landand Embankments in BangladeshMusahaq Ali, Zakaria Hossain and Shigeko Haruyama
Graduate School of Bioresources, Mie University, Japan
ABSTRACTThe influence of tropical cyclones on flood height and damages of land and embankments in Bangladesh have beenstudied based on the field visits to cyclone site, collected data and information on damages of land and embankments
during field visits. Review of technical literature on tropical cyclones obtained from recent publications was conducted
for a comparative study of this research. An analysis of death of peoples, damages of lands, embankments andproperties, flood surge heights, wind speed and radius of the severe storm were demonstrated. It is concluded that thesouth side of Bangladesh was not only the affected region but also the other parts of the country such as Sirajgonj
district and Jamuna River were greatly affected due to the tropical cyclones in Bangladesh. The Bay of Bengal wasparticularly vulnerable to tropical cyclones where at least sixteen major cyclones including four severe cyclonic storms
occurred that killed over 400 thousand people after the independence of Bangladesh..
Keywords: Tropical Cyclones, Severity, Agriculture, Land, Flood height, Wind speed, Death
1. INTRODUCTION
The tropical cyclones occurred in the year of 1970 (Bhola),
1991 (Tropical), 1999 (Orissa) and 2007 (Sidr) are four
severe cyclones along with over 16 major cyclones occurredin Bangladesh (Hossain et al). The 1970 Bhola cyclone was adevastating cyclone that struck on November 12, 1970 (Islam,2006). It was the deadliest tropical cyclone ever recorded, andone of the deadliest natural disasters in modern times. Morethan 250,000 people lost their lives in the storm, primarily asa result of the storm surge that flooded much of the low-lying
islands of the Ganges Delta. This cyclone was the sixth
cyclonic storm of the 1970 North Indian Ocean cyclone
season, and was also the most powerful, reaching a strength
equivalent to a Category 3 hurricane. This cyclone formed
over the central Bay of Bengal on November 8 and traveled
north, intensifying as it did so. It reached its peak with windsof 185 km/h (115 mph) on November 12, and made landfall
on the coast of East Pakistan that night. The storm surge
devastated many of the offshore islands, wiping out villages
and destroying crops throughout the region. The city of Thana,Tazumuddin, was the most severely affected, with over 45%
of the population of 167,000 killed by the storm (Paul andRahman, 2006).
The cyclone which struck Bangladesh on the night of 29-30,
April, 1991 was particularly severe causing widespread
damage, killing 138,882 people (Bern, et al 1993). There has
been massive damage to life line systems as well as private
properties. Total loss has been estimated at US$2.07 billion
dollars for all sectors. Cyclone on 18th October, ’99 hit on theeastern cost of India along the Bay of Bengal affecting thecoastal districts of Ganjam, Puri, Jagatsinghpur, Khurda,

Gajapati and Balasore. Subsequently, the State was hit by aSuper Cyclone on 29th October, ’99 with winds of more than
250 km/h, tidal waves rising 20 feet high and torrential rains.
The Super Cyclone and its aftermath caused "severe" damage
in the districts of Jagatsinghpur, Balsore, Cuttack, Puri,
Nayagarh, Jajpur Kendrapada, Bhadrak and Khurda and"moderate" damage in the districts of Mayurbhanj,
Dhenkanal and Keonjhar (BWDB. 1998). In 2007,
Bangladesh suffered a natural disaster like the cyclone Sidr ofNovember 15, an unusually powerful storm that triggered
giant waves up to 30ft (7m) high and killed more than 10,000
people in the south western coastal belt of Bangladesh
covering the districts of Bagerhat, Barisal, Patuakhali,
Pirozepur, Khulna and Satkhira. Even five days after thecalamity, in village after village along the battered roads ofthe coastal districts, the survivors of the storm, their cheeks
hollow and eyes sunken from hunger wait in vain for relief tocome. True, natural disasters are a universal reality. But inmodern times, it is the preparedness that counts. Damage canbe minimized, rehabilitation can be effective, but only if anation has an organized disaster management plan. Unhappily,
Bangladesh does not. Accompanied by 260 km/h ravaging
winds and sea surge as high as 30ft that swept about 20 km
inland, a grade 5 hurricane, the highest level possible,
devastated 23 districts in the south western part of
Bangladesh at midnight on November 15, 2007 (EU, 1998).
The country is yet to fully repair many of the damages done toher infrastructure due to devastating flood occurred in thesame year just three months ago (Hossain and Sakai, 2008).
Agricultural infrastructures such as farm structures, irrigation
structures, dairy, poultry, fisheries, shelter, sanitation,
drinking water, electricity supplies, transportation services
including both land and water transports are in great danger inthe region of Bay of Bengal where natural disasters occurred881

frequently. Considering the above facts, the damages ofagricultural land, flood heights and death of people alongwith other factor are focused in this paper.
2. PATH OF CYCLONES AND SEVERIETIESThe path of the four severe cyclones along with the physicalfacts of Bangladesh is shown in Fig.2. It should be noted that
Bangladesh extends between 21o and 27o North latitude and88o and 92.5o East longitude. The Bay of Bengal is in thesouth side of the country. The total area is 144,000 sq km andsize of population is around 15 million (2007). Per capita
income is around US$ 550, one of the lowest in the world.
The coastal land of Bangladesh (710 km long) is of recent
origin formed out of the process of sedimentation. Most partsof the area are, therefore, low lying which can be subject toinundation even under ordinary circumstances of tides. A
tidal surge accompanied by a cyclone storm makes thesituation alarming which is further exacerbated by thetriangular shape of the Bay of Bengal (Fig.1).

Bay and adjoining Andaman Sea on 23 April. Finally, thecyclone of hurricane intensity crossed the Chittagong coast alittle north of Chittagong at 2 a. m. of 30 April, 1991 thatkilled some 143,000 people in Bangladesh. The aspects of thedetection of the cyclone, its monitoring and prediction andweakness of warning and special weather bulletins have beenreflected upon. Most of the worst affected thanas are either
off-shore islands or coastal thanas. The less affected thanas
are mostly located inside and further from the coast. The 2007
cyclone Sidr smashed into the country's southern coastline
late on Thursday midnight of November 15 with 250 km/h
(155 m/h) winds that whipped up a five meter tidal surge andswept about 20 km inland, a grade 5 hurricanes, the highest
level possible, devastated 23 districts in the south westernpart of Bangladesh. Gigantic walls of water smashed into thecoastline, washing away everything in their path. The
backwash dragged hundreds of people into the sea, bodies
were towards the shoreline -- twisted, bloated and broken -washing up from the overflowing creeks and ponds aroundthe villages. Death and its nauseating stench wereeverywhere.
3. EFFECT ON WIND, FLOOD AND LAND
In order to understand the severity of storm of the tropical
cyclones in Bangladesh, a comparative study of the maximum
wind speed and storm surge of major cyclones those struck in
Bay of Bengal during the last four decades is depicted in
Figs.2 and 3. The high power of the storm was mainly due tothe higher wind speed which is apparent from Fig.2 for all thecases after 1970. In 1970 that is known as Bhola cyclone,though the wind speed was not as high as of the recent
cyclones that can be found in 1991, 1998 and 2007, however,the number of casualties was severe at that time. The windspeed of Bhola cyclone was nearly 227 km/h whereas theothers were over 230 km/h such as 257, 237 and 247 km/h forthe cyclones occurred in the year of 1991, 1999 and 2007,
respectively. The storm surge, on the other hand, wasrecorded as 7.8, 8.8, 8.1 and 8.4 m for the cyclones occurredin the year of 1991, 1999 and 2007, respectively, as shown in
Fig.3. The radius of the major cyclones is shown in Fig.4.

882

300
250
200
150
100
50
0
19
70
19
73
19
7
19 6
78
19
81
19
83
19
85
19
8
19 8
91
19
94
19
96
19
98
19
9
20 9
02
20
04
20
07

The wide shallow continental shelf is conducive toamplification of surges causing wide spread flooding. The
country has been subjected to frequent natural disasters inmany forms, particularly cyclonic storms and tidal surges.
From 1970 to 2007, four major cyclone storms and tidal
surges have been reported. These indicate that Bangladesh isprone to frequent destructive tropical cyclones associatedwith tidal surge. The low-lying coastal areas are particularly
vulnerable, thus placing these population, infrastructure,
agriculture, livestock and economic development in ahigh-risk situation. During the 1991 cyclone, the cyclonicstorm was detected as a low pressure area over the Southeast

Maximum wind speed (km/h)

Fig.1 Paths of 4 extremely severe cyclones

Year of occurance

Fig.2 Wind Speed during the major cyclones in Bangladesh

Year of occurance

30

Death of number ofpeopleX104

80
70
60
50
40
30
20
10
0

Death

Damage

6

25

5

20

4

15

3

10

2

5

1

0
1970

19
7
19 0
7
19 3
7
19 6
7
19 8
8
19 1
8
19 3
8
19 5
8
19 8
9
19 1
9
19 4
9
19 6
9
19 8
9
20 9
0
20 2
0
20 4
07

Radius of maximum windspeed (km)Fig.3 Flood surge height during the major cyclones in
Bangladesh

Damages in billion US$

10
9
8
7
6
5
4
3
2
1
0
19
7
19 0
7
19 3
7
19 6
7
19 8
8
19 1
8
19 3
8
19 5
8
19 8
9
19 1
9
19 4
9
19 6
9
19 8
9
20 9
0
20 2
0
20 4
07

Storm surge (m)The damage in 1999 and 2007 is noted as 4.2 and 5.3 billion
US dollar. A large number of boats and smaller ships ran
aground. Bangladesh Navy and Air Force were also heavily
hit. BNS Isha Khan Naval Base was flooded, with heavy
damage to the ships. Most of the fighter planes of the Air
Force were also damaged. Comparison between the death ofpeople and the damages in Fig.5 in terms of money in
US$ during the past four major cyclones notices that the
number of death of people is getting to decrease although thedamages of the properties increases in the recent years. This
clearly indicates the awareness and preparedness of thepeople in the coastal region, improvement of weather
broadcasting system and countermeasures taken by thegovernment of the country. However, these measures could
not be considered enough yet and more action need to betaken to minimize the death and damages of the peoples, their

0
1980

1990

2000

2010

Year of occuranceYear of occurance

Fig.5 Number of death of people and damages of severe
cyclonic storms in Bangladesh

Fig.4 Maximum radius of major cyclones

4.

EFFECT ON PEOPLE AND EMBANKMENT

The number of death of people and the damages due to major
cyclones in Bangladesh are given in Fig.6 as the line graphs
for the sake of clarification of the ferocious devastation
brought to Bangladesh by the cyclones in just 37 years.
Although the wind speed and tidal surge was lower for the1970 Bhola cyclone, than the recent ones, the number ofdeath of people was extremely higher at that time. It wasestimated that more than 250000 people was died during the1970 Bhola cyclone. It was saying that this was the deadliest
one in the earth in its history. On the other hand, the 1991
cyclone was also other deadliest tropical cyclones on recordthat struck greater Chittagong with strong winds and storm
surge, killing at least 140000 people and leaving as many as
one crore people homeless and destroying 10 lakh homes.
The damage of property was estimated at 1.8 billion US dollar
in 1991.

homes, cattle, agriculture, forestry and fisheries, as well as its
environment, bio-diversity, Sundarbans (world largest
mangrove) and wildlife.
5. DISCUSSION
There was widespread damage to transport andcommunications networks. Rural roads, and many of theembankments protecting such roads, were extensively
damaged. Most of the road damage was associated with thetidal surges in coastal areas. Large uprooted trees onroadsides also account for some of the damages, as trees were
uprooted segments of tarmac or earthen roads became
cracked or fragmented. Damage to transport infrastructure incoastal and inland waterways had occurred. Numerous
ferries, and associated landing and loading areas damaged. In
more than a few cases, the storm surge was so strong thatmedium to large sized ferries were actually lifted clear out of
the water and beached on neighboring land.

883

6. CONCLUSIONS
Nineteen ninety one’s cyclone killed at least 138,000 people,
and the 1970 cyclone in Bhola, which left as many as above
250,000 people dead and is considered the deadliest cyclone,
and one of the worst natural disasters, in human history. Sidr
2007 is another recent deadliest one causing thousands ofpeople dead, millions of acres of cropland washed down by
the sweeping ocean-surge, one third of Sundarban, a world
natural heritage, utterly torn down, and substantial
infrastructural damages have caused combined losses ofassets and agricultural production. About 3,931 km longcoastal embankment to protect coastal land from inundation
by tidal waves and storm-surges, and drainage channels of
total length of 4,774 km have so far been constructed, but lackof maintenance has rendered them almost ineffective. The
results reported in this paper can be used as the preparatory
information about flood height and scale of effect on land andpeople of Bangladesh.
7. REFERENCES[1] Hossain MZ, Islam MT, Sakai T and Ishida M. Impact ofTropical Cyclones on Rural Infrastructures in
Bangladesh, Agricultural Engineering International: the
CIGR Ejournal. Invited Overview No. 2, Vol. X. April,
2008[2] Islam, T. 2006. Integrated Approach to Cyclone Wind
Analysis and disaster planning for the Benladesh coast,
Ph.D. Dissertation, Texas Tech University, December.
[3] Paul, A., and Rahman, M. 2006. Cyclone Mitigation
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Sandwip and Hatia Islands, Coastal Management, 34,
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[4] Bern, C., Sniezek, J., Mathbor, G. M., Siddiqi, M. S.,
Ronsmans, C., Chowdhury, A. M., Choudhury, A. E.,
Islam, K., Bennish, M., and Noji, E. 1993. Risk factors
for mortality in the Bangladesh cyclone of 1991, Bulletinof the World Health Organization, 71(1), pp.73-78.
[5] BWDB. 1998. DHV, Meghna Estuary Study, Draft
Master Plan, Volume 1, Main Report, for BWDB,
Dhaka.
[6] EU (European Community). 1998. Cyclone Shelter
Preparatory Study (CPSP), Stage I, Feasibility Study,
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[7] Hossain, M.Z. and Sakai, T. Extent and Severity of Flood
Embankments in Bangladesh, Agricultural Engineering
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[8] UN (United Nation), 2007. Cyclone Sidr United Nations
Rapid Initial Assessment Report, With a Focus on 9
Worst Affected Districts, 22 November.

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Authors Index
A

Ali Sobhanmanesh, Malaysia

A. A. Mirghasemi, Iran

Alireza sadeghabadi, Iran

A. Abdul Rahman, India

Alkroosh, Nikraz, Australia

A. Elahi, Iran

Aly Ahmed, Egypt

A. Ghalandarzadeh, IranAmin Falamaki, Iran

A. Gholampour, IranAmin Naeimabadi, IranA. Iizuka, Japan

Aniruddha Senguptar, India

A. M. Al-Mahbashi, Saudi Arabia

Anna Grzyb, France

A. Rajendra Prasad, India

Anna Maria Pellegrino, Italy

A. Rajendra Prasad, India

Antoine Guimond-Barrett, France

A. Tohidi, Iran

Ardalan Akbari Hamed, Iran

A. Abdul Rahman, India

Arifin Abdu, Malaysia

A. Dhanalakshmi, India

Arvin Abdolmaleki, Iran

A. Haghollahi, IranAsskar Janalizade Choobbasti, IranA. Iizuka, Japan

Atrchian mohammad reza, Iran

A. Kanazawa, Japan

Atsuko Nonomura, Japan

A. Kumar, Australia

Azlan Bin Adnan, Malaysia

ASM Abdul Awal

Azza M. Elleboudy, Egypt

Abbas Ghalandarzadeh, IranAbbasali Taghavi, Iran

B

Abdoullah Namdar, Malaysia

B. Gatmiri, IranAfshin Asadi, Malaysia

B. Maher, Australia

Ahaiti Reheman, Japan

B. Rafieey, IRAN

Ahmad Fahimifar, Iran

B.G. Muhammad, Malaysia

Ahmed A. Mohamed, Egypt.

Bipul Hawlader, Canada

Ahmed H. Abdul Kareem, Iraq

Bour Moslem, Iran

Ahmed Hassan, Egypt

Bujang bin Kim Huat, Malaysia

Ahmed M. Eltohamy, Egypt.
Ahmed Rusdy Towfek, Egypt.

C

Akhila Manne, India

C. Behnia, Iran

Akihiro Hashimoto, Japan

C. S. Chen, Malaysia

Akira Kikuchi, Malaysia

Carrillo-Acevedo, Peru

Akram Y. Al-Sa’aty, Iraq

Carrillo-Gil, Peru

Al Serif M. Abd Alazez, Egypt.

Chaminda Gallage, Australia

Alain Le Kouby, France

Charrismanagara Arry, Australia

Ali Ghozat, Iran

Choo Chung Siung, Malaysia

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Hara, T., JapanD

Hasan A. Kamal, Kuwait

D. Elmi, Iran

Hasan Sharafi, Iran

D. Wanatowski, United Kingdom

Hassan J. Karam, Kuwait

D.E.L. Ong, Malaysia

Hayano, K.

Daghigh younes, IranHijiri Hashimoto, Japan

Dominic Ong E. L, Malaysia

Hirochika Hayashi, Japan

Dygku Salma Awg Ismail, Malaysia

Hirochika Hayashi, Japan
Hiroo Shiojiri, JapanF

Hiroshi Masuya, Japan

Fabien Szymkiewicz, France

Hiroyasu Ohtsu, Japan

Faisal Ali, Malaysia

Hongyu Qin, Australia

Fakher Ali, Iran

Hooman Abadi, Malaysia

Farshad Fayyaz Jahani, Iran

Hossein Moayedi, Iran

Farzad Allahyari, Iran

Hossein Tavallali, Iran

Fathi M. O. Mohamed, Canada

Hyun-Ku Park, Korea

Ferry Fatnanta, Indonesia
Frederic Pellet, France

I

Fukawa, H. , Japan

Ideris Bin Zakaria, Malaysia

Fumitake Nishimura, Japan

Iman Ashayeri, Iran

G

J

G.R. Reddyr, India

J. M. Kate, India

Gauhar Sabih, Pakistan

J. Rajaraman, India

Ghanbari namin hamed, Iran

J. Santhosh, India

Gholami Meade, Iran

J. Shamshuddin, Malaysia

Gholamreza Darvishi, Iran

Jahangir Khazaie, Iran

Gholamreza Shoaei, Iran

Jahangir Khazaie, Iran

Gi-Dae Oh, Korea

Jan Jaremski, Poland

Gratchev Ivan, Australia

Janaka J. Kumara, Japan
Jia He, USA

H

Jian Chu, USA

H. Elahi, Iran

Jitsangiam P., Australia

H. S. Sagady, Sudan

John N. Mukabi, Kenya

Hakime Rabbanifar, Iran

Jun Sugawara, Australia

Hamed A. Keykha, Malaysia
Hamed Bavanpouri, IranK

Hamidreza Saba, Kuwait

K. Kawai, Japan886

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K. Kawai, Japan

M. P. Bhorkar, India

K. Oda, Japan

M. Shafieyzadeh, Iran

K. Yamamoto, Iran

M.S LEE, Japan

K.Thiruvenkatasamy, India

M.S. Jaafar, Malaysia

Kamran Akhtar, Pakistan

Mahdi Ali Bolandi, Iran

Kaoru Kobayashi, Japan

Mahdy Khari, Malaysia

Kazuhiro Oda, Japan

Mahmoud Dhaybi, France

Kazuo Fujita, Japan

Mahnoosh Biglari, IranKeizo Ugai, Egypt

Mahshid Mirzaaghaee, Iran

Kenichiro Nakarai, Japan

Majid Mirzaei, Malaysia

Kenji Hotta, Japan

Man-Kwon Choi, Japan

Khairul Anuar Bin Kassim, Malaysia

Masaaki Ishida, Japan

Khazaie Jahangir, Iran

Masaaki Kondo, Japan

Khosro Ebrahimkhani, Iran

Masahiko Katoh, Japan

Kiarash Ashtari, Iran

Md. Bellal Hossain, Japan

Kimitoshi Hayano, Japan

Md. Eahya Al Huda, India

Koichi Onohara, Japan

Md. Iftekharuzzaman, Canada

Kota Sasaki, Japan

Md. Toriqul Islam, Japan

Kyoichi Okamoto, Japan

Md. Zakaria Hossain, Japan
Mehdi Arab Ameri, IranL

Mehdi Arab Ameri, IranL G Tham, Hong Kong

Mehdi Mokhberi, Australia

L.R. Stace, United Kingdom

Michko Masuda, Japan

Lee Min Lee, Malaysia

Minatour Yasser, Iran

Lee Sieng Kai, Malaysia

Minoru YAMANAKA, Japan

Leonardo Schippa, Italy

Minsun Lee, JapanLim Lion Yee, Malaysia

Mitsuhiro Inoue, Japan
Mohamad Sharifipour, Iran

M

Mohamad T. Al-Layla, Iraq

M H. Sadagiani, Iran

Mohammad Ahmadi, Iran.

M. A. Dafalla, Saudi Arabia

Mohammad Arab Ameri, Iran

M. Abdoli, Iran

Mohammad Arab Ameri, Iran

M. Akiyama, Japan

Mohammad Gheitury, Iran.

M. Dianah, Malaysia

Mohammad Mahdi Roozbahani , Malaysia

M. Fathalikhani, Iran

Mohammad Moghadaripour, Iran

M. Hira, Japan

Mohammad Reza Atrchian Iran

M. Mirshekari, Iran

Mohammad Reza Atrechian, Iran

M. Moradi, Iran

Mohammad S. Pakbazy, IRAN

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Second International Conference on Geotechnique, Construction Materials and Environment,
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Mohammad Sharifipour, Iran

Omar Mustafa, Iraq

Mohammed K. Fekheraldin, Iraq

Osama M. A. Daoud, Sudan

Mohd Raihan Taha, Malaysia

Osama M. A. Daoud, Sudan

Morikawa, Y. Japan
Morteza Ala, IranP

Mosayeb Heshmati, IranP. Egodawatta, Australia

Mosayeb Heshmati, Iran.

P. Sethu Raja, India

Moshe Livneh, Israel

Philippe Reiffsteck, France

Mosleh Al Shamrani1, Saudi Arabia

Poona Abadi, Malaysia

Mostafa Yousefi Rad, IranPoona Abadi, Malaysia

Muhamad Faiz, Malaysia

Purwana, Y.M., Australia

Muhamad Ghaituri, IranMusa Mutah, Malaysia

Q

Musahaq Ali, Japan

Qutayba Nazar Al-Saffar, Iraq

Mu’azu Mohammed Abdullahi, Malaysia
R
N

R. Ziaie Moayed, Iran

N. Aramaki, Japan

R.Sanjeev, India

N. Rostam Afshar, Iran

Raid R. Al-Omari, Iraq

N.Z. Mohd Yunus, Malaysia

Ramin Doostmohammadi, IranNaoto Takashita, Japan

Ramli Nazir, Malaysia

Naoya Toyota, Japan

Reza Ziaie Moayed, IranNarges Janalizadeh, Malaysia

Rouzbeh Abadi, Malaysia

Navid Khayat, Iran

Rouzbeh Abadi, Malaysia

Navid Shahnazi, IranNazri Bin Ali, Malaysia

S

Neelima Satyam, India

S. Fakhretaha, IranNg Wei Soon, Malaysia

S. Jayakody, Australia

Nik Muhamad Majid, Malaysia

S. Kawasaki, Japan

Nikraz H., Australia

S. Muhammad Jamil, Pakistan

Noam A. Livneh, Israel

S. Nomura, JapanNor Eman Ismail, Malaysia

S. Ohno, Japan

Noriyuki Yasufuku, Japan

S. Raei, Iran

Nuri al-Mohamadi, Jordan

S. Sakamoto, Japan

Nurly Gofar, Malaysia

S. Tsurumi, japan

Nurmunira Binti Muhammad, Malaysia

S.M. Abdul Awal, Malaysia
S.M. Binesh, Iran
S.Nishiyama, JapanO

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S.Ohno, Japan

Sunil Kumar, India

S.Tsurumi, Japan

Surendra Singh, India

S.Usha, India

Syawal SATIBI, Indonesia

Sabzi Zahra, IranSai

TK. Vanapalli1, Canada

Salam J. Bash Al-Maliky, Iraq

T. Negami, Japan

Salman Hanifi, Iran

T. Souliyavong, Australia

Salman hanifi, Iran

T. Y. Elkady, Saudi Arabia

Samaneh Kargar, Iran

T.Kikuchi, Japan

Satoru Kawasaki, Japan

T.Yano, Japan

Satoru Kawasaki, Japan

Taghizadeh Maryam, Iran

Satoshi MASAKI, Japan

Takaaki Tsukada, Japan

Satoshi Nishimoto, Japan

Takamitsu Kajisa, Japan

Satoshi Nishimoto, Japan

Takayuki Isoda, Japan

Satoshi Nishimoto, Japan

Takayuki Kawai, Japan

Sawada, K., Japan

Takehana, K.

Seok Yoon, Korea

Takeshi SATO, Japan

Seung-Rae Lee, Korea

Tam Sy HO, Japan

Seyed Ali Farjam, Australia

Tan Jui Siang, Malaysia

Seyed Ali Hashemi, Iran

Tanaka, S.

Seyed Salman Musavi, IranTaro Uchimura, Japan

Shaharin Ibrahim, Malaysia

Tetsuya Suzuki, Japan

Shahrbanoo Rezanejad Farahmand, IranToshihiro Morii, Japan

Shigeko Haruyama, Japan

Toshimitsu Komatsu, Japan

Shinsuke Matsumoto, Japan

Toshinori Sakai, Japan

Shinya Inazumi, Japan

Tschuschke W. , Poland

Shiran Jayakody, Australia

Tsuguhiro Nonaka, Japan

Shokouhi Ali, Australia

Tsuji, S., Japan

Shotaro Kitamura, Japan
Shuichi HASEGAWA, JapanV

Shushi Sato, Japan

V. Phommachanh, Japan

Siamak Pahlevanzadeh, Australia

V. Rashidian, Iran

Siavash Zamiran, Kuwait

Viktor Stabnikov, USA

Sina Kazemian, IranVivi Anggraini, Malaysia

Siti Noor Linda Taib, Malaysia

Volodymyr Ivanov, USA

Soewignjo Agus Nugroho, Indonesia
Sonia Fanelli, France

W

Srijit Bandyopadhyayr, India

W.W. Ng, Hong Kong

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Wei Dong Guo, Australia

List of Countries

Wierzbicki J. , Poland
Australia
X

CanadaXu Kai, Hong Kong

Egypt
FranceY

Hong Kong

Y. Hayashi, Japan

India

Y. Sugiyama, Japan

Indonesia

Y. Takayama, JapanIranY.Takayama, Japan

Iraq

Yasuo Tanaka, Malaysia

IsraelYoshida, M., Japan

Italy

Yoshihiro Kimura, JapanJapan

Yoshio Suematsu, Japan

JordanYousefi Kebria Daryoush, IranKeniyaYousefi Kebria Daryoush, Iran

Korea

Yousefi kebria Daryoush, Iran

Saudi Arabia

Yousry M. Mowafe, Egypt.

KuwaitYuki Hasegawa, Japan

Malaysia

Yuki Yokoyama, Japan

Nigeria

Yuuki Shigekuni, Japan

Pakistan
Peru

Z

PolandZakaria Hossain, Japan

Singapore

Zeinab Bakhshipour, Malaysia

Sudan

Zieaoddin Shoaei, Iran

United Kingdom
United States of America

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