GEOMAT2011 MIE
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTi

ii

PROCEEDINGS OF FIRST INTERNATIONAL CONFERENCE - GEOMAT2011
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT MIE JAPAN, 21-23
NOVEMBER, 2011

Geotechnique, Construction
Materials and Environment
Volume 1
Edited by

Md. Zakaria Hossain
Department of Environmental Science and Technology
Mie University

Toshinori Sakai
Department of Environmental Science and Technology
Mie University

THE GEOMATE INTERNATIONAL SOCIETY, MIE 2011iii

Copyright @ 2011 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrate and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: geomate@gi-j.com
www.gi-j.com

ISBN Number: 978-4-9905958-0-7 C3051

Printed in Japan
iv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Table of Contents

Prefacexi
xii

Organization

Geotechnique
Developments of geosynthetic-reinforced soil technology in Japan: retaining walls and bridge
abutments
Fumio Tatsuoka

3

Seepage-deformation coupled numerical analysis of unsaturated river embankment using anelasto-plastic model
F. Oka, S.Kimoto and R.Kato

15

Progressive failure of shallow anchor in dense sand evaluated by 2D and 3D finite element
analysis and model tests
T. Sakai and M. Rokonuzzaman

23

Design of single pile foundations using the mechanics of unsaturated soils
Vanapalli S. and Taylan Z.N.

31

Problematic soil: in search for solution
Bujang B.K. Huat

39

Characterization and modeling of various aspects of pre-failure deformation of clayey
geomaterials – fundamental theories and analyses
J.N. Mukabi

47

The maintenance of ground anchors in Nippon Expressway
Yuu Fujiwara and Toshinori Sakai

57

Stress distribution in khon kaen loess under spread footing
Pairoj Yodsa-nga and Watcharin Gasaluck

61

The effect of sheet pile length on the capacity of improved pile foundation on sand undervertical and moment loading
Tuan Van Nguyen and Pongsakorn Punrattanasin

65

Investigation of a reliable reinforcing method for embankment ground
H. M. Shahin, T. Nakai, M. Kikumoto, R. Sakai, Y. Yoshida, I. Saito, N. Nagao and K. Toda

71

Evaluation of stamps and latex shell influence on stabilometer testing results, modeled withsimulia abaqus
Dmitriy Volik

77

Characterizing the interaction of geomat with fine grained black cotton soils
Wekesa S., Mukabi J. , Okado J., Kogi S., Ndeda M. and Ogallo J.

81

Soil liquefaction susceptibility zoning due to earthquake using GIS and geotechnical data
Bhuyan Md. Habibullah, Rama Mohan Pokhrel, Jiro Kuwano and Shinya Tachibana

87v

Evaluation on the results of multistage shear test
Hormdee D. and Kaikeerati N.

93

Verification of effects of countermeasures for seepage of river levees by on-site monitoring of
the water level of levees
Hiroyuki Masuyama, Yukiko Saito, Masanori Ishihara, Tetsuya Sasaki and Hirotoshi Mori

97

Influence of decreasing of soil suction on unconfined compressive strength for bentonite
Tomoyoshi Nishimura103

The relationship between soil suction and the maximum unsaturated undrained shear
strengths of compacted khon kaen soil
R. Nuntasarn and W. Wannakul

107

Interpretation of coaxial and non-coaxial strain on directional variation of undrained strength ofclay
J.Rajaraman and K. Thiruvenkatasamy

113

FEM analysis of innovative shallow foundation system for reactive soils
Mostafa A. Ismail and Mohamed A. Shahin

117

Delay times of elastic waves at discontinuities in laminated specimens
Yota Togashi, Kazuo Tani, Tetsuji Okada and Hiroaki Sato

123

Geotechnical investigation of valleys using lightweight dynamic cone penetration test forlandslide risk assessment
Takashi Hanaoka, Takashi Tsuchida, Shoichi Kawabata and Shota Nakagawa

129

Study on individual landslide risk assessment of natural valleys and slopes during heavy rain
based on geotechnical investigation and analysis
Shouichi Kawabata, Takashi Tsuchida, Takashi Hanaoka and Shouta Nakagawa

135

Analyses of piping under foundation of weirs in different ground density by FEM
Okajima Kenji141

Shear deformation and failure of sandy slope according to pore pressure generation due torainfall infiltration
Katsuo Sasahara and Naoki Sakai

147

Load settlement relationships of circular footing considering dilatancy characteristics of sand
Yusuke Tomita, Tatsuaki Nishigata, Takeshi Masui and Shintaro Yao

153

Thermally modifying bentonite for construction industry
Abdoullah Namdar

159

Mesh-free analysis of beam on elastic foundation
S. M. Binesh

163

Frictional Performance Of Infilled I-Blocks With Geosynthetic Inclusions
Faisal Hj Ali, Md. Zahidul Islam Bhuiyan, Firas A. Salman and Siau Lian San

167

Varying ground water level to minimise liquefaction hazards in urban areas
Heitor A., Oka F., Kimoto S. and Higo Y.

173

Development of a new type reinforced soil wall
Hara T., Tsuji S., Yoshida M., Ito S. and Sawada K.

177

Development of design charts for reinforced embankment using excel spreadsheet slope
stability analysis
Ito Kazuki, Md. Zakaria Hossain and Toshinori Sakai

183

vi

Numerical simulation of 2D crack growth with frictional contact in brittle materials
Kai Zhang, Jing-cai Jiang and Qing Yang

187

Interface behaviour of basalt geosynthetic with sand in direct shear mode
M.B. Hossain, M.Z. Hossain and T. Sakai

191

Effect of soil parameter uncertainty on seismic response of buried segmented pipeline
Chaminda Gallage, Pathmanathan Rajeev and Jayantha Kodikara

197

The sorption capacity of cu by sc-soil from batch and column tests
Arwut Yimtae, Pongsakorn Punrattanasin and Noppadol Sangiumsak

203

Construction materials
Green construction: the role of palm oil fuel ash in concrete
A.S.M. Abdul Awal

211

Behavior of collapsible loessic soil after interparticle cementation
Arrua Pedro, Aiassa Gonzalo, Eberhardt Marcelo and Claverie Eliana

221

A study on hardened soil brick using liquid type and powder type hardening agent
Jin Sung Ki., Jung Seung Yong. and Kim Tae Ho.

227

The effect of salt content in soil on the sorption capacity of heavy metals
Lerd Kerdchaiyaphum and Pongsakorn Punrattanasin

233

Manufacture of Housing Construction Materials with Jute Polymer Composite (Jutin)
Mubarak A. Khan, Fahmida Parvin, Jahid M M Islam and Md. Zakaria Hossain

239

Evaluation of strength of granulated construction waste sludge with quick lime and its
application to soil structure
Makoto Kobayashi, Hiroaki Nakamura and Hideaki Mizuno

245

Analysis of enhanced strength and deformation resistance of some tropical geomaterials
through application of in-situ based stabilization techniques
Kogi S., Ndeda M., Mukabi J.N. and Wekesa S.

251

Further developments in the swelling model of expansive clays under ASTM 4546 and CBR
testing
Moshe Livneh

257

The effect of clay, lime and rice husk ash contents on the sorption capacity of Cu, Ni and Zn by
Soils
Pongsakorn Punrattanasin and Noppadol Sanggiumsak

263

The effect of temperature on the sorption capacity of copper by soils
Noppadol Sangiumsak and Pongsakorn Punrattanasin

269

Experimental Investigation for Nitrate Reduction Using Iron Powder in Porous MediaHaruka Shimada, Kazuya Inoue, Ikko Ihara, Katsutoshi Suzuki and Tsutomu Tanaka

275

Using spatial moments in conjunction with image processing to estimate macrodispersion in
stratified porous formations
Katsutoshi Suzuki, Kazuya Inoue, Tsutomu Tanaka and Akira Kobayashi

281

The novel technology of desertification control and ecological restoration with an Hydrolysis
Polyurethane(W-OH)
Weimin Gao, Zhiren Wu, Zhishen Wu, Kentaro Iwashita, Hirondo Inagaki and Caiqian Yang

vii

287

Migration of radioactive materials through porous media: theory and a solution for an unsteadystate problem
Khaled I. Hamza

293

Fundamental study on ecosystem support canal using porous concrete
Masuma Yoshihiro, Hasegawa Yuki, Md.Rokunuzzaman, Yamada Toshio,Kassai Hirofumi and
Sato Shushi

299

Alternative employment of crushed shell particles in capillary barrier of soil
Satoru Nakafusa, Kaoru Kobayashi, Toshihiro Morii,Tomoyoshi Nishimura

305

A novel statistical approach for investigating the significant factors that influence theperformance score of the Turkish method
Kshama Sundar Roy, Kamrul Islam and Md. Shams Arafat

311

Ultimate strength and fatigue life of concrete structural elements reinforced with carbon fiber
sheet
Hidenori Tanaka and Yutaka Toi

317

Development of functional carbon nanotubes -asphalt composites
Tatsuo Shirakawa, Akio Tada and Noriyasu Okazaki

323

Hydraulic character of estimation method on roughness coefficient of concrete canal
Koichiro Otagaki, Koji Sai,Kenichi Fujisawa, Shinsuke Matsumoto and Shushi Sato329

Na-bentonite as rock containment barrier against heavy metals and acid leachates
Angelica Naka, Takeshi Katsumi, Toru Inui, Giancarlo Flores and Takehiro Ohta

333

The comparison of monotonic behaviors of two different calcareous sands
R. Rezvani, H. Shahnazari, H. Salehzadeh and Y. Dehnavi

337

Soaking effect on strength and performance of fine grained soil stabilized with recycled
gypsumAly Ahmed, Keizo Ugai and Taira Jinno

343

Removal of Zinc in soils using natural zeolite
Wiparat Nisapai , Arthit Neramittagapong and Sutasinee Neramittagapong

349

Heat control of pavement surface temperature using oyster shell lime
Satoru Ishiguro and Masayoshi Yamanaka

353

Buckling behavior of composite triangular plates
T.Farsadi, P.Amani and E.Heydarnia

357

Durability of cement treated highway materials
Jiraroth Sukolrat, Apisit Klummeng and Sekchai Anuvechsirikiat

365

Porous geopolymer concrete
Tawatchai Tho-in, Vanchai Sata and Prinya Chindaprasirt

371

Design and Fabrication of Composite Technology for Earth Slope Protection
Md. Toriqul Islam, Md. Zakaria Hossain and Masaaki Ishida375

Study of slump testing for hydraulic plastic grout
Hiroyasu Ishii, Kenichi Horikoshi and Masaki Kitazume381

Behavioral comparison of coarse aggregate from various quarry sites of pakistan, effect on
hardened concrete, economics and environment
Faisal Malik and S. Muhammad Jamil

viii

387

Methods for execution of concrete piles in corrosive and destructive marine environments,
based on study of the Persian Gulf marine installations
Nasser Arafati, Abdol Hossein Fazli and Seyed Mehdi Mousavi

393

Fundamental study on particle size distribution of coarse materials by image analysis
G.H.A.J.J. Kumara, K. Hayano and K. Ogiwara

399

Environment
Calibration and simulation of runoff models for developing watershed respond knowledge at
tropical area
Arien Heryansyah and Zulkifli Yusop

407

Effect of geological succession on macrophyte and microbiota in aquifer ecosystem in urban
coastal zone
Kazuhito Murakami and Akiko Inoue-Koham

411

Bioclean and liquid biofertilizers a new way to the green area
Sittisak Uparivong

417

Massive point cloud acquisition on a soil surface using a stereo camera
Masafumi Nakagawa, Anna Nakanishi, Hirotaka Endo and Kenta Ochiai

421

Estimation of earthquake ground motion in Padang, Indonesia
Rusnardi Rahmat Putra, J. Kiyono and Y. Ono

425

Groundwater contamination due to irrigation of treated sewage effluent in the werribee delta
Hiroyuki Ii, Masanobu Taniguchi, Ataru Satoh, Graeme Allinson, Matt Kitching, George
Croatto and Bruce Shelley

431

Process of invading the alien plant species into the river ecosystem
Michiko Masuda, Yumie Ito and Fumitake Nishimura

437

Recent meandering pattern of the Irrawaddy, Myanmar
Mayumi Matsumoto and Shigeko Haruyama

443

Physical and chemical properties of tsunami deposit in northeast area in Fukushima prefecture
after Tohoku-Kanto earthquake
Tomonori Fujikawa, Hiromu Okazawa, Takahiko Nakamura, Yasushi Takeuchi and Masaharu
Komamura

447

Changes of bottom sediment caused by Construction of the Airport Island in the Ise Bay, Japan.
Maki Oyagi, Motohiro Kawase and Akihiko Yagi

453

Proposal of simple measurement method for evaporation rate by using oxygen isotopic ratio in
the inawashiro lake
Satoshi Miyahara and Hiroyuki Ii

457

Landslide susceptibility mapping using logistic regression model with neighborhood analysis:
a case study in mizunami city
Wang Liangjie, Sawada Kazuhide and Moriguchi Shuji463

The evaluation of the river environment and water quality by image analysis in the YamatoRiver Basin in Japan
Masanobu Taniguchi, Hiroyuki Ii and Tatemasa Hirata

469

Distribution of oxygen and hydrogen stable isotopic ratios and estimation of the spring waterorigin in the Kushiro Moor
Natsuki Kawabe and Hiroyuki Ii

473

ix

A debris flow and its risk analysis related to the 2008 Wenchuan earthquake
Yange Li, Guangqi Chen, Chuan Tang and Lu Zheng

479

Impact analysis of water price reform of Zhangye, China
Atsushi Koike, Zhongmin Xu, Kang Wang and Bunei Itoga

485

Post earthquake rapid inspection planning for bridges in western Kentucky
Norimitsu Koike and Issam E. Harik

491

Organophosphoric acid triesters from landfill sites and sewage plants in Japan
Haruki Shimazu

497

Ground water pollution around the Oyachi-Heizu district of Yokkaichi city
Yukimasa Takemoto, Masaaki Takahashi, Kayoko Awaya, Mikihiro Ioka and Peng Guo

501

Characteristics of damage to structures induced by the tsunami of the 2011 east Japan megaearthquakeÖmer Aydan, Hisataka Tano and Takahiro Iwatate

505

Characteristics of strong motions induced by the 2011 east Japan mega earthquake with anemphasis on Tokyo bay area
Takahiro Iwatate, and Ömer Aydan

511

Inquiry of the value of parks in the characteristics and use of park planning through urban
revival planning projects in Maebashi city
Shinya Tsukada, Tetsuo Morita and Akira Yuzawa

517

A study on evaluation of quality of life in consideration of water/green environment
Tetsuo Morita, Yoshihito Kogure, Hiroshi Sugita, Tsuyoshi Baba, Shinya Tsukada and Naoki
Miyazato

521

Changes and issues in green space planning in the tokyo metropolitan area: focusing on the"capital region plan"
Tetsuo Morita, Yoshihide Nakagawa, Akinori Morimoto, Masateru Maruyama and Yoshimi
Hosokawa

527

Application of traffic simulation in road policy assessment accompanied by railway elevation
Maekawa Takuya and Hiroshi Inouye

533

Impact estimation of land cover changes of recharge area on spring discharge in marshy
environment
Masaaki Kondo and Takamitsu Kajisa

541

Numerical study on characteristics of levee breach flood disasters in alluvial floodplain
Md. Serazul Islam and Tetsuro Tsujimoto

547

Purification system of ocean sludge by activating microorganisms
Kyoichi Okamoto, Kenji Hotta, Takeshi Toyama and Hideki Kohno

553

Author index

559

Volume 2563

xFirst International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Preface

The First International Conference on Geotechnique, Construction Materials and Environment
GEOMAT2011 is organized in Center Palace Miyako Hotel, Tsu city, Mie Japan in conjunction with Japanese
Geotechnical Society (JGS), Japanese Society of Irrigation Drainage and Rural Engineering (JSIDRE), Mie
University, JCK Comp. Ltd and Glorious International. It aims to provide with great opportunities to sharecommon interests on geo-engineering, construction materials, environmental issues, water resources, andearthquake and tsunami disasters. The key objective of this conference is to promote interdisciplinary
research from various regions of the globe. On Friday 11 March at 14:46 Japan Standard Time, the north eastof Japan was severely damaged by the tragic earthquake and tsunami. The conference is dedicated to the
tragic Tohoku-Kanto earthquake and tsunami disasters.
The conference has 3 major themes with 17 specific themes including● Advances in Composite Materials● Computational Mechanics● Foundation and Retaining Walls● Slope Stability● Soil Dynamics● Soil-Structure Interaction● Pavement Technology● Tunnels and Anchors● Site Investigation and Rehabilitation● Ecology and Land Development● Water Resources Planning● Environmental Management● Public Health and Rehabilitation● Earthquake and Tsunami Issues● Safety and Reliability● Geo-Hazard Mitigation● Case History and Practical Experience
The proceedings contain 7 keynote papers along with 115 technical papers from 22 countries. The technicalpapers are selected from the vast number of contributions submitted, after review of the abstracts. The final
papers included in the proceedings have been peer reviewed rigorously and revised as necessary by theauthors. We are grateful to the authors of the contributed papers for their resourceful papers and their help inmaintaining the high assessment of the papers and the co-operation in complying with the requirements of theeditor and the reviewers. We wish to express our sincere thanks to the Organizing Committee Members,
National Advisory Committee Members and International Advisory Committee Members for their valuable
supports. We acknowledge the support of Japanese Geotechnical Society (JGS), Japanese Society of
Irrigation Drainage and Rural Engineering (JSIDRE), Mie University, JCK Comp. Ltd and Glorious
International.

Prof. Dr. Toshinori Sakai
Conference Chairman

Dr. Md. Zakaria Hossain
Conference SecretaryxiFirst International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Organization
Scientific Committees:
Conference Chairman: Prof. Dr. Toshinori Sakai, Mie University, Japan
Conference Honorary Chairman: E/Prof. Dr. Sohji Inoue, Mie University, Japan
Conference Organizing Committee:
Prof. Dr. Takamitsu Kajisa, Mie University, Japan
Prof. Dr. Ishiguro Satoru, Mie University, Japan
Prof. Dr. Sigeko Haruyama, Mie University, Japan
A/Prof. Dr. Atsushi Ishii, Mie University, Japan
A/Prof. Dr. Maasaki Kondo, Mie University, Japan
A/Prof. Dr. Kenji Okajima, Mie University, Japan
A/Prof. Dr. Ryoei Ito, Mie University, Japan
National Advisory Committee:
Prof. Dr. Fumio Tatsuoka, Tokyo University of Science, Japan
Prof. Dr. Kazunori Uchida, Kobe University, Japan
Prof. Dr. Fusao Oka, Kyoto University, Japan
Prof. Dr. Jing-Cai Jiang, University of Tokushima, Japan
Prof. Dr. Akira Murakami, Kyoto University, Japan
E/Prof. Dr. Tadatsugu Tanaka, University of Tokyo, Japan
A/Prof. Dr. Kingshuk Roy, Nihon University, Japan
A/Prof. Dr. Md. Shahin Hossain, Nagoya Institute of Technology, Japan
A/Prof. Dr. Sayuri Kimoto, Kyoto University, Japan
Engr. Hirondo Inagali, President, JCK Com. Ltd. Japan
A/Prof. Dr. Inui Toru, Kyoto University, Japan
A/Prof. Dr. Fujisawa Kazunori, Okayama University, Japan
A/Prof. Dr. Inoue Kazuya, Kobe University, Japan
International Advisory Committee:
Prof Sai Vanapalli, University of Ottawa, Canada
Prof. Dr. Mounir Bouassida, National School of Engineering of Tunis
Prof. Dr. Lilia Robles Austriaco, Angles University Foundation, Philippines
Prof. Dr. A.S.M. Abdul Awal, University Technology Malaysia, Malaysia
Prof. Dr. Muhammad Ibn Ibrahimy, International Islamic University, Malaysia
Prof. Dr. Zahurul Islam, Bangladesh University of Engineering and Technology, Bangladesh
Prof. Dr. Bujang B.K. Huat, University Putra Malaysia, Malaysia
Prof. Dr. Nemy Banthia, University of British Columbia, Canada
Prof. Dr. Ian Jefferson, University of Birmingham, UK
Prof. Dr. John Bolander, Univesity of California, USA
Prof. Dr. Shamshul Chowdhury, Roosevelt University, USA
Prof. Dr. Isabel Pinto, University of Coimbra, Portugal
A/Prof. Mark Jaksa, University of Adelaide, Australia
A/Prof. Dr. William Sebudde Kaggwa, University of Adelaide, Australia
Dr. John N. Mukabi, President of Kenya Geotechncal Society, Kenyaxii

Conference Secretariat:
Dr. Zakaria Hossain, Associate Professor
Division of Environmental Science and Technology
Graduate School of Bioresources
Mie University, 1577 Kurima Machiya-cho
Tsu-city, Mie 514-8507, Japan
E-mail: zakaria@bio.mie-u.ac.jp
Tel: +81-59-231-9578、Fax: +81-59-231-9591
Editorial Committee:
Md. Zakaria Hossain
Md. Bellal Hossain
Toshinori Sakai
Md. Toriqul Islam
Executive Committee:
Toshinori Sakai
Md. Zakaria Hossain
Md. Bellal Hossain
Kazuki Ito
Md. Toriqul Islam
Muhammad Ibn Ibrahimy
Ms. Emi Ueno
Ms. Mayumi MatsumotoxiiiGEOMAT2011 MIE
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTi

ii

PROCEEDINGS OF FIRST INTERNATIONAL CONFERENCE - GEOMAT2011
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT MIE JAPAN, 21-23
NOVEMBER, 2011

Geotechnique, Construction
Materials and Environment
Volume 2
Edited by

Md. Zakaria Hossain
Department of Environmental Science and Technology
Mie University

Toshinori Sakai
Department of Environmental Science and Technology
Mie University

THE GEOMATE INTERNATIONAL SOCIETY, MIE 2011iii

Copyright @ 2011 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrate and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: geomate@gi-j.com
www.gi-j.com

ISBN Number: 978-4-9905958-0-7 C3051

Printed in Japan
iv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Table of Contents

Prefacevi
vii

Organization
Characterization and modeling of various aspects of pre-failure deformation of clayey
geomaterials – applications in modeling
J.N. Mukabi and M.Z. Hossain

563

An emergency medical care information system for fetal monitoring
Muhammad Ibn Ibrahimy

573

Effects of gypsum on the swelling and dispersion behaviour of fine grained soils from Sudan
Abdul Karim, M. Zein and Ahmed M. Elsharief

577

Comparison of bearing capacity equations for vertical central loading
Muhammad Awais Shafique and Tanvir Iqbal Qayyum

583

Pull-out resistance of selected native plants in Hong Kong
Leung F.T.Y., Yan W.M., Tham L.G. and Hau B.C.H.

589

Inland navigation in carrying injured and sick disaster victims in case of a massive earthquake
in the Tokyo metropolitan area
Kazuya Egami, Keiichi Noto, Wataru Miyazaki, Kazukiyo Yamamoto and Takeo Kondo

593

Model calculation and geotechnical interpretation of the correlation co-efficient between finite
and incremental strain path of a formation
J.Rajaraman and K. Thiruvenkatasamy

597

Geocell reinforced bed subjected to simulated traffic loads
P. Sharifi, and S.N. Moghaddas Tafreshi

601

Effects of gradation of various sand deposits of Pakistan on strength of hardened concrete
Gauhar Sabih and S. Muhammad Jamil607

Strengthening of landslide retaining pile wall after geodetic monitoring
Frangov G., Zayakova H., and Hamova M.

613

Response of tapered piles under lateral harmonic vibrations
Ahmad Dehghanpoor and Mahmoud Ghazavi

617

Investigation of ground movement in soil-nailed walls using 1g models and PIV method
H. Mahdavi, and M. Hajialilue-Bonab

623
629

Author Indexv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Preface

The First International Conference on Geotechnique, Construction Materials and Environment
GEOMAT2011 is organized in Center Palace Miyako Hotel, Tsu city, Mie Japan in conjunction with Japanese
Geotechnical Society (JGS), Japanese Society of Irrigation Drainage and Rural Engineering (JSIDRE), Mie
University, JCK Comp. Ltd and Glorious International. It aims to provide with great opportunities to sharecommon interests on geo-engineering, construction materials, environmental issues, water resources, andearthquake and tsunami disasters. The key objective of this conference is to promote interdisciplinary
research from various regions of the globe. On Friday 11 March at 14:46 Japan Standard Time, the north eastof Japan was severely damaged by the tragic earthquake and tsunami. The conference is dedicated to the
tragic Tohoku-Kanto earthquake and tsunami disasters.
The conference has 3 major themes with 17 specific themes including● Advances in Composite Materials● Computational Mechanics● Foundation and Retaining Walls● Slope Stability● Soil Dynamics● Soil-Structure Interaction● Pavement Technology● Tunnels and Anchors● Site Investigation and Rehabilitation● Ecology and Land Development● Water Resources Planning● Environmental Management● Public Health and Rehabilitation● Earthquake and Tsunami Issues● Safety and Reliability● Geo-Hazard Mitigation● Case History and Practical Experience
The proceedings contain 7 keynote papers along with 115 technical papers from 22 countries. The technicalpapers are selected from the vast number of contributions submitted, after review of the abstracts. The final
papers included in the proceedings have been peer reviewed rigorously and revised as necessary by theauthors. We are grateful to the authors of the contributed papers for their resourceful papers and their help inmaintaining the high assessment of the papers and the co-operation in complying with the requirements of theeditor and the reviewers. We wish to express our sincere thanks to the Organizing Committee Members,
National Advisory Committee Members and International Advisory Committee Members for their valuable
supports. We acknowledge the support of Japanese Geotechnical Society (JGS), Japanese Society of
Irrigation Drainage and Rural Engineering (JSIDRE), Mie University, JCK Comp. Ltd and Glorious
International.

Prof. Dr. Toshinori Sakai
Conference Chairman

Dr. Md. Zakaria Hossain
Conference Secretaryvi

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Organization
Scientific Committees:
Conference Chairman: Prof. Dr. Toshinori Sakai, Mie University, Japan
Conference Honorary Chairman: E/Prof. Dr. Sohji Inoue, Mie University, Japan
Conference Organizing Committee:
Prof. Dr. Takamitsu Kajisa, Mie University, Japan
Prof. Dr. Ishiguro Satoru, Mie University, Japan
Prof. Dr. Sigeko Haruyama, Mie University, Japan
A/Prof. Dr. Atsushi Ishii, Mie University, Japan
A/Prof. Dr. Maasaki Kondo, Mie University, Japan
A/Prof. Dr. Kenji Okajima, Mie University, Japan
A/Prof. Dr. Ryoei Ito, Mie University, Japan
National Advisory Committee:
Prof. Dr. Fumio Tatsuoka, Tokyo University of Science, Japan
Prof. Dr. Kazunori Uchida, Kobe University, Japan
Prof. Dr. Fusao Oka, Kyoto University, Japan
Prof. Dr. Jing-Cai Jiang, University of Tokushima, Japan
Prof. Dr. Akira Murakami, Kyoto University, Japan
E/Prof. Dr. Tadatsugu Tanaka, University of Tokyo, Japan
A/Prof. Dr. Kingshuk Roy, Nihon University, Japan
A/Prof. Dr. Md. Shahin Hossain, Nagoya Institute of Technology, Japan
A/Prof. Dr. Sayuri Kimoto, Kyoto University, Japan
Engr. Hirondo Inagali, President, JCK Com. Ltd. Japan
A/Prof. Dr. Inui Toru, Kyoto University, Japan
A/Prof. Dr. Fujisawa Kazunori, Okayama University, Japan
A/Prof. Dr. Inoue Kazuya, Kobe University, Japan
International Advisory Committee:
Prof Sai Vanapalli, University of Ottawa, Canada
Prof. Dr. Mounir Bouassida, National School of Engineering of Tunis
Prof. Dr. Lilia Robles Austriaco, Angles University Foundation, Philippines
Prof. Dr. A.S.M. Abdul Awal, University Technology Malaysia, Malaysia
Prof. Dr. Muhammad Ibn Ibrahimy, International Islamic University, Malaysia
Prof. Dr. Zahurul Islam, Bangladesh University of Engineering and Technology, Bangladesh
Prof. Dr. Bujang B.K. Huat, University Putra Malaysia, Malaysia
Prof. Dr. Nemy Banthia, University of British Columbia, Canada
Prof. Dr. Ian Jefferson, University of Birmingham, UK
Prof. Dr. John Bolander, Univesity of California, USA
Prof. Dr. Shamshul Chowdhury, Roosevelt University, USA
Prof. Dr. Isabel Pinto, University of Coimbra, Portugal
A/Prof. Mark Jaksa, University of Adelaide, Australia
A/Prof. Dr. William Sebudde Kaggwa, University of Adelaide, Australia
Dr. John N. Mukabi, President of Kenya Geotechncal Society, Kenyavii

Conference Secretariat:
Dr. Zakaria Hossain, Associate Professor
Division of Environmental Science and Technology
Graduate School of Bioresources
Mie University, 1577 Kurima Machiya-cho
Tsu-city, Mie 514-8507, Japan
E-mail: zakaria@bio.mie-u.ac.jp
Tel: +81-59-231-9578、Fax: +81-59-231-9591
Editorial Committee:
Md. Zakaria Hossain
Md. Bellal Hossain
Toshinori Sakai
Md. Toriqul Islam
Executive Committee:
Toshinori Sakai
Md. Zakaria Hossain
Md. Bellal Hossain
Kazuki Ito
Md. Toriqul Islam
Muhammad Ibn Ibrahimy
Ms. Emi Ueno
Ms. Mayumi MatsumotoviiiGEOMAT2011 MIE
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTi

ii

PROCEEDINGS OF FIRST INTERNATIONAL CONFERENCE - GEOMAT2011
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT MIE JAPAN, 21-23
NOVEMBER, 2011

Geotechnique, Construction
Materials and Environment
Volume 1
Edited by

Md. Zakaria Hossain
Department of Environmental Science and Technology
Mie University

Toshinori Sakai
Department of Environmental Science and Technology
Mie University

THE GEOMATE INTERNATIONAL SOCIETY, MIE 2011iii

Copyright @ 2011 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrate and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: geomate@gi-j.com
www.gi-j.com

ISBN Number: 978-4-9905958-0-7 C3051

Printed in Japan
iv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Table of Contents

Prefacexi
xii

Organization

Geotechnique
Developments of geosynthetic-reinforced soil technology in Japan: retaining walls and bridge
abutments
Fumio Tatsuoka

3

Seepage-deformation coupled numerical analysis of unsaturated river embankment using anelasto-plastic model
F. Oka, S.Kimoto and R.Kato

15

Progressive failure of shallow anchor in dense sand evaluated by 2D and 3D finite element
analysis and model tests
T. Sakai and M. Rokonuzzaman

23

Design of single pile foundations using the mechanics of unsaturated soils
Vanapalli S. and Taylan Z.N.

31

Problematic soil: in search for solution
Bujang B.K. Huat

39

Characterization and modeling of various aspects of pre-failure deformation of clayey
geomaterials – fundamental theories and analyses
J.N. Mukabi

47

The maintenance of ground anchors in Nippon Expressway
Yuu Fujiwara and Toshinori Sakai

57

Stress distribution in khon kaen loess under spread footing
Pairoj Yodsa-nga and Watcharin Gasaluck

61

The effect of sheet pile length on the capacity of improved pile foundation on sand undervertical and moment loading
Tuan Van Nguyen and Pongsakorn Punrattanasin

65

Investigation of a reliable reinforcing method for embankment ground
H. M. Shahin, T. Nakai, M. Kikumoto, R. Sakai, Y. Yoshida, I. Saito, N. Nagao and K. Toda

71

Evaluation of stamps and latex shell influence on stabilometer testing results, modeled withsimulia abaqus
Dmitriy Volik

77

Characterizing the interaction of geomat with fine grained black cotton soils
Wekesa S., Mukabi J. , Okado J., Kogi S., Ndeda M. and Ogallo J.

81

Soil liquefaction susceptibility zoning due to earthquake using GIS and geotechnical data
Bhuyan Md. Habibullah, Rama Mohan Pokhrel, Jiro Kuwano and Shinya Tachibana

87v

Evaluation on the results of multistage shear test
Hormdee D. and Kaikeerati N.

93

Verification of effects of countermeasures for seepage of river levees by on-site monitoring of
the water level of levees
Hiroyuki Masuyama, Yukiko Saito, Masanori Ishihara, Tetsuya Sasaki and Hirotoshi Mori

97

Influence of decreasing of soil suction on unconfined compressive strength for bentonite
Tomoyoshi Nishimura103

The relationship between soil suction and the maximum unsaturated undrained shear
strengths of compacted khon kaen soil
R. Nuntasarn and W. Wannakul

107

Interpretation of coaxial and non-coaxial strain on directional variation of undrained strength ofclay
J.Rajaraman and K. Thiruvenkatasamy

113

FEM analysis of innovative shallow foundation system for reactive soils
Mostafa A. Ismail and Mohamed A. Shahin

117

Delay times of elastic waves at discontinuities in laminated specimens
Yota Togashi, Kazuo Tani, Tetsuji Okada and Hiroaki Sato

123

Geotechnical investigation of valleys using lightweight dynamic cone penetration test forlandslide risk assessment
Takashi Hanaoka, Takashi Tsuchida, Shoichi Kawabata and Shota Nakagawa

129

Study on individual landslide risk assessment of natural valleys and slopes during heavy rain
based on geotechnical investigation and analysis
Shouichi Kawabata, Takashi Tsuchida, Takashi Hanaoka and Shouta Nakagawa

135

Analyses of piping under foundation of weirs in different ground density by FEM
Okajima Kenji141

Shear deformation and failure of sandy slope according to pore pressure generation due torainfall infiltration
Katsuo Sasahara and Naoki Sakai

147

Load settlement relationships of circular footing considering dilatancy characteristics of sand
Yusuke Tomita, Tatsuaki Nishigata, Takeshi Masui and Shintaro Yao

153

Thermally modifying bentonite for construction industry
Abdoullah Namdar

159

Mesh-free analysis of beam on elastic foundation
S. M. Binesh

163

Frictional Performance Of Infilled I-Blocks With Geosynthetic Inclusions
Faisal Hj Ali, Md. Zahidul Islam Bhuiyan, Firas A. Salman and Siau Lian San

167

Varying ground water level to minimise liquefaction hazards in urban areas
Heitor A., Oka F., Kimoto S. and Higo Y.

173

Development of a new type reinforced soil wall
Hara T., Tsuji S., Yoshida M., Ito S. and Sawada K.

177

Development of design charts for reinforced embankment using excel spreadsheet slope
stability analysis
Ito Kazuki, Md. Zakaria Hossain and Toshinori Sakai

183

vi

Numerical simulation of 2D crack growth with frictional contact in brittle materials
Kai Zhang, Jing-cai Jiang and Qing Yang

187

Interface behaviour of basalt geosynthetic with sand in direct shear mode
M.B. Hossain, M.Z. Hossain and T. Sakai

191

Effect of soil parameter uncertainty on seismic response of buried segmented pipeline
Chaminda Gallage, Pathmanathan Rajeev and Jayantha Kodikara

197

The sorption capacity of cu by sc-soil from batch and column tests
Arwut Yimtae, Pongsakorn Punrattanasin and Noppadol Sangiumsak

203

Construction materials
Green construction: the role of palm oil fuel ash in concrete
A.S.M. Abdul Awal

211

Behavior of collapsible loessic soil after interparticle cementation
Arrua Pedro, Aiassa Gonzalo, Eberhardt Marcelo and Claverie Eliana

221

A study on hardened soil brick using liquid type and powder type hardening agent
Jin Sung Ki., Jung Seung Yong. and Kim Tae Ho.

227

The effect of salt content in soil on the sorption capacity of heavy metals
Lerd Kerdchaiyaphum and Pongsakorn Punrattanasin

233

Manufacture of Housing Construction Materials with Jute Polymer Composite (Jutin)
Mubarak A. Khan, Fahmida Parvin, Jahid M M Islam and Md. Zakaria Hossain

239

Evaluation of strength of granulated construction waste sludge with quick lime and its
application to soil structure
Makoto Kobayashi, Hiroaki Nakamura and Hideaki Mizuno

245

Analysis of enhanced strength and deformation resistance of some tropical geomaterials
through application of in-situ based stabilization techniques
Kogi S., Ndeda M., Mukabi J.N. and Wekesa S.

251

Further developments in the swelling model of expansive clays under ASTM 4546 and CBR
testing
Moshe Livneh

257

The effect of clay, lime and rice husk ash contents on the sorption capacity of Cu, Ni and Zn by
Soils
Pongsakorn Punrattanasin and Noppadol Sanggiumsak

263

The effect of temperature on the sorption capacity of copper by soils
Noppadol Sangiumsak and Pongsakorn Punrattanasin

269

Experimental Investigation for Nitrate Reduction Using Iron Powder in Porous MediaHaruka Shimada, Kazuya Inoue, Ikko Ihara, Katsutoshi Suzuki and Tsutomu Tanaka

275

Using spatial moments in conjunction with image processing to estimate macrodispersion in
stratified porous formations
Katsutoshi Suzuki, Kazuya Inoue, Tsutomu Tanaka and Akira Kobayashi

281

The novel technology of desertification control and ecological restoration with an Hydrolysis
Polyurethane(W-OH)
Weimin Gao, Zhiren Wu, Zhishen Wu, Kentaro Iwashita, Hirondo Inagaki and Caiqian Yang

vii

287

Migration of radioactive materials through porous media: theory and a solution for an unsteadystate problem
Khaled I. Hamza

293

Fundamental study on ecosystem support canal using porous concrete
Masuma Yoshihiro, Hasegawa Yuki, Md.Rokunuzzaman, Yamada Toshio,Kassai Hirofumi and
Sato Shushi

299

Alternative employment of crushed shell particles in capillary barrier of soil
Satoru Nakafusa, Kaoru Kobayashi, Toshihiro Morii,Tomoyoshi Nishimura

305

A novel statistical approach for investigating the significant factors that influence theperformance score of the Turkish method
Kshama Sundar Roy, Kamrul Islam and Md. Shams Arafat

311

Ultimate strength and fatigue life of concrete structural elements reinforced with carbon fiber
sheet
Hidenori Tanaka and Yutaka Toi

317

Development of functional carbon nanotubes -asphalt composites
Tatsuo Shirakawa, Akio Tada and Noriyasu Okazaki

323

Hydraulic character of estimation method on roughness coefficient of concrete canal
Koichiro Otagaki, Koji Sai,Kenichi Fujisawa, Shinsuke Matsumoto and Shushi Sato329

Na-bentonite as rock containment barrier against heavy metals and acid leachates
Angelica Naka, Takeshi Katsumi, Toru Inui, Giancarlo Flores and Takehiro Ohta

333

The comparison of monotonic behaviors of two different calcareous sands
R. Rezvani, H. Shahnazari, H. Salehzadeh and Y. Dehnavi

337

Soaking effect on strength and performance of fine grained soil stabilized with recycled
gypsumAly Ahmed, Keizo Ugai and Taira Jinno

343

Removal of Zinc in soils using natural zeolite
Wiparat Nisapai , Arthit Neramittagapong and Sutasinee Neramittagapong

349

Heat control of pavement surface temperature using oyster shell lime
Satoru Ishiguro and Masayoshi Yamanaka

353

Buckling behavior of composite triangular plates
T.Farsadi, P.Amani and E.Heydarnia

357

Durability of cement treated highway materials
Jiraroth Sukolrat, Apisit Klummeng and Sekchai Anuvechsirikiat

365

Porous geopolymer concrete
Tawatchai Tho-in, Vanchai Sata and Prinya Chindaprasirt

371

Design and Fabrication of Composite Technology for Earth Slope Protection
Md. Toriqul Islam, Md. Zakaria Hossain and Masaaki Ishida375

Study of slump testing for hydraulic plastic grout
Hiroyasu Ishii, Kenichi Horikoshi and Masaki Kitazume381

Behavioral comparison of coarse aggregate from various quarry sites of pakistan, effect on
hardened concrete, economics and environment
Faisal Malik and S. Muhammad Jamil

viii

387

Methods for execution of concrete piles in corrosive and destructive marine environments,
based on study of the Persian Gulf marine installations
Nasser Arafati, Abdol Hossein Fazli and Seyed Mehdi Mousavi

393

Fundamental study on particle size distribution of coarse materials by image analysis
G.H.A.J.J. Kumara, K. Hayano and K. Ogiwara

399

Environment
Calibration and simulation of runoff models for developing watershed respond knowledge at
tropical area
Arien Heryansyah and Zulkifli Yusop

407

Effect of geological succession on macrophyte and microbiota in aquifer ecosystem in urban
coastal zone
Kazuhito Murakami and Akiko Inoue-Koham

411

Bioclean and liquid biofertilizers a new way to the green area
Sittisak Uparivong

417

Massive point cloud acquisition on a soil surface using a stereo camera
Masafumi Nakagawa, Anna Nakanishi, Hirotaka Endo and Kenta Ochiai

421

Estimation of earthquake ground motion in Padang, Indonesia
Rusnardi Rahmat Putra, J. Kiyono and Y. Ono

425

Groundwater contamination due to irrigation of treated sewage effluent in the werribee delta
Hiroyuki Ii, Masanobu Taniguchi, Ataru Satoh, Graeme Allinson, Matt Kitching, George
Croatto and Bruce Shelley

431

Process of invading the alien plant species into the river ecosystem
Michiko Masuda, Yumie Ito and Fumitake Nishimura

437

Recent meandering pattern of the Irrawaddy, Myanmar
Mayumi Matsumoto and Shigeko Haruyama

443

Physical and chemical properties of tsunami deposit in northeast area in Fukushima prefecture
after Tohoku-Kanto earthquake
Tomonori Fujikawa, Hiromu Okazawa, Takahiko Nakamura, Yasushi Takeuchi and Masaharu
Komamura

447

Changes of bottom sediment caused by Construction of the Airport Island in the Ise Bay, Japan.
Maki Oyagi, Motohiro Kawase and Akihiko Yagi

453

Proposal of simple measurement method for evaporation rate by using oxygen isotopic ratio in
the inawashiro lake
Satoshi Miyahara and Hiroyuki Ii

457

Landslide susceptibility mapping using logistic regression model with neighborhood analysis:
a case study in mizunami city
Wang Liangjie, Sawada Kazuhide and Moriguchi Shuji463

The evaluation of the river environment and water quality by image analysis in the YamatoRiver Basin in Japan
Masanobu Taniguchi, Hiroyuki Ii and Tatemasa Hirata

469

Distribution of oxygen and hydrogen stable isotopic ratios and estimation of the spring waterorigin in the Kushiro Moor
Natsuki Kawabe and Hiroyuki Ii

473

ix

A debris flow and its risk analysis related to the 2008 Wenchuan earthquake
Yange Li, Guangqi Chen, Chuan Tang and Lu Zheng

479

Impact analysis of water price reform of Zhangye, China
Atsushi Koike, Zhongmin Xu, Kang Wang and Bunei Itoga

485

Post earthquake rapid inspection planning for bridges in western Kentucky
Norimitsu Koike and Issam E. Harik

491

Organophosphoric acid triesters from landfill sites and sewage plants in Japan
Haruki Shimazu

497

Ground water pollution around the Oyachi-Heizu district of Yokkaichi city
Yukimasa Takemoto, Masaaki Takahashi, Kayoko Awaya, Mikihiro Ioka and Peng Guo

501

Characteristics of damage to structures induced by the tsunami of the 2011 east Japan megaearthquakeÖmer Aydan, Hisataka Tano and Takahiro Iwatate

505

Characteristics of strong motions induced by the 2011 east Japan mega earthquake with anemphasis on Tokyo bay area
Takahiro Iwatate, and Ömer Aydan

511

Inquiry of the value of parks in the characteristics and use of park planning through urban
revival planning projects in Maebashi city
Shinya Tsukada, Tetsuo Morita and Akira Yuzawa

517

A study on evaluation of quality of life in consideration of water/green environment
Tetsuo Morita, Yoshihito Kogure, Hiroshi Sugita, Tsuyoshi Baba, Shinya Tsukada and Naoki
Miyazato

521

Changes and issues in green space planning in the tokyo metropolitan area: focusing on the"capital region plan"
Tetsuo Morita, Yoshihide Nakagawa, Akinori Morimoto, Masateru Maruyama and Yoshimi
Hosokawa

527

Application of traffic simulation in road policy assessment accompanied by railway elevation
Maekawa Takuya and Hiroshi Inouye

533

Impact estimation of land cover changes of recharge area on spring discharge in marshy
environment
Masaaki Kondo and Takamitsu Kajisa

541

Numerical study on characteristics of levee breach flood disasters in alluvial floodplain
Md. Serazul Islam and Tetsuro Tsujimoto

547

Purification system of ocean sludge by activating microorganisms
Kyoichi Okamoto, Kenji Hotta, Takeshi Toyama and Hideki Kohno

553

Author index

559

Volume 2563

xFirst International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Preface

The First International Conference on Geotechnique, Construction Materials and Environment
GEOMAT2011 is organized in Center Palace Miyako Hotel, Tsu city, Mie Japan in conjunction with Japanese
Geotechnical Society (JGS), Japanese Society of Irrigation Drainage and Rural Engineering (JSIDRE), Mie
University, JCK Comp. Ltd and Glorious International. It aims to provide with great opportunities to sharecommon interests on geo-engineering, construction materials, environmental issues, water resources, andearthquake and tsunami disasters. The key objective of this conference is to promote interdisciplinary
research from various regions of the globe. On Friday 11 March at 14:46 Japan Standard Time, the north eastof Japan was severely damaged by the tragic earthquake and tsunami. The conference is dedicated to the
tragic Tohoku-Kanto earthquake and tsunami disasters.
The conference has 3 major themes with 17 specific themes including● Advances in Composite Materials● Computational Mechanics● Foundation and Retaining Walls● Slope Stability● Soil Dynamics● Soil-Structure Interaction● Pavement Technology● Tunnels and Anchors● Site Investigation and Rehabilitation● Ecology and Land Development● Water Resources Planning● Environmental Management● Public Health and Rehabilitation● Earthquake and Tsunami Issues● Safety and Reliability● Geo-Hazard Mitigation● Case History and Practical Experience
The proceedings contain 7 keynote papers along with 115 technical papers from 22 countries. The technicalpapers are selected from the vast number of contributions submitted, after review of the abstracts. The final
papers included in the proceedings have been peer reviewed rigorously and revised as necessary by theauthors. We are grateful to the authors of the contributed papers for their resourceful papers and their help inmaintaining the high assessment of the papers and the co-operation in complying with the requirements of theeditor and the reviewers. We wish to express our sincere thanks to the Organizing Committee Members,
National Advisory Committee Members and International Advisory Committee Members for their valuable
supports. We acknowledge the support of Japanese Geotechnical Society (JGS), Japanese Society of
Irrigation Drainage and Rural Engineering (JSIDRE), Mie University, JCK Comp. Ltd and Glorious
International.

Prof. Dr. Toshinori Sakai
Conference Chairman

Dr. Md. Zakaria Hossain
Conference SecretaryxiFirst International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Organization
Scientific Committees:
Conference Chairman: Prof. Dr. Toshinori Sakai, Mie University, Japan
Conference Honorary Chairman: E/Prof. Dr. Sohji Inoue, Mie University, Japan
Conference Organizing Committee:
Prof. Dr. Takamitsu Kajisa, Mie University, Japan
Prof. Dr. Ishiguro Satoru, Mie University, Japan
Prof. Dr. Sigeko Haruyama, Mie University, Japan
A/Prof. Dr. Atsushi Ishii, Mie University, Japan
A/Prof. Dr. Maasaki Kondo, Mie University, Japan
A/Prof. Dr. Kenji Okajima, Mie University, Japan
A/Prof. Dr. Ryoei Ito, Mie University, Japan
National Advisory Committee:
Prof. Dr. Fumio Tatsuoka, Tokyo University of Science, Japan
Prof. Dr. Kazunori Uchida, Kobe University, Japan
Prof. Dr. Fusao Oka, Kyoto University, Japan
Prof. Dr. Jing-Cai Jiang, University of Tokushima, Japan
Prof. Dr. Akira Murakami, Kyoto University, Japan
E/Prof. Dr. Tadatsugu Tanaka, University of Tokyo, Japan
A/Prof. Dr. Kingshuk Roy, Nihon University, Japan
A/Prof. Dr. Md. Shahin Hossain, Nagoya Institute of Technology, Japan
A/Prof. Dr. Sayuri Kimoto, Kyoto University, Japan
Engr. Hirondo Inagali, President, JCK Com. Ltd. Japan
A/Prof. Dr. Inui Toru, Kyoto University, Japan
A/Prof. Dr. Fujisawa Kazunori, Okayama University, Japan
A/Prof. Dr. Inoue Kazuya, Kobe University, Japan
International Advisory Committee:
Prof Sai Vanapalli, University of Ottawa, Canada
Prof. Dr. Mounir Bouassida, National School of Engineering of Tunis
Prof. Dr. Lilia Robles Austriaco, Angles University Foundation, Philippines
Prof. Dr. A.S.M. Abdul Awal, University Technology Malaysia, Malaysia
Prof. Dr. Muhammad Ibn Ibrahimy, International Islamic University, Malaysia
Prof. Dr. Zahurul Islam, Bangladesh University of Engineering and Technology, Bangladesh
Prof. Dr. Bujang B.K. Huat, University Putra Malaysia, Malaysia
Prof. Dr. Nemy Banthia, University of British Columbia, Canada
Prof. Dr. Ian Jefferson, University of Birmingham, UK
Prof. Dr. John Bolander, Univesity of California, USA
Prof. Dr. Shamshul Chowdhury, Roosevelt University, USA
Prof. Dr. Isabel Pinto, University of Coimbra, Portugal
A/Prof. Mark Jaksa, University of Adelaide, Australia
A/Prof. Dr. William Sebudde Kaggwa, University of Adelaide, Australia
Dr. John N. Mukabi, President of Kenya Geotechncal Society, Kenyaxii

Conference Secretariat:
Dr. Zakaria Hossain, Associate Professor
Division of Environmental Science and Technology
Graduate School of Bioresources
Mie University, 1577 Kurima Machiya-cho
Tsu-city, Mie 514-8507, Japan
E-mail: zakaria@bio.mie-u.ac.jp
Tel: +81-59-231-9578、Fax: +81-59-231-9591
Editorial Committee:
Md. Zakaria Hossain
Md. Bellal Hossain
Toshinori Sakai
Md. Toriqul Islam
Executive Committee:
Toshinori Sakai
Md. Zakaria Hossain
Md. Bellal Hossain
Kazuki Ito
Md. Toriqul Islam
Muhammad Ibn Ibrahimy
Ms. Emi Ueno
Ms. Mayumi MatsumotoxiiiGeotechniqueFirst International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Developments of Geosynthetic-Reinforced Soil Technology in Japan: Retaining
Walls and Bridge Abutments
Tatsuoka, F., Department of Civil Engineering, Tokyo University of Sciencesoils to facilitate both drainage and tensile reinforcing of
the backfill. The latter makes possible the use of
low-quality on-site soil as the backfill if necessary.

ABSTRACT:The
constructionofpermanent
geosynthetic-reinforced soil (GRS) retaining walls (RWs) with afull-height rigid facing for railways, including high-speed train lines,
started about twenty years ago in Japan. The total length of this typeof GRS RW is now more than 125 km, replacing traditional
gravity-type RWs or cantilever reinforced concrete RWs andsteel-reinforced soil RWs. Many were also constructed replacing
traditional type RWs and embankments that collapsed during recent
earthquakes, heavy rainfalls and storms. By taking advantages ofthis technology, a number of bridge abutments withgeosynthetic-reinforced backfill were constructed. The latest version,
called the GRS integral bridge, comprises a continuous girder
integrated to a pair of RC facing with the backfill reinforced withgeosynthetic reinforcement layers firmly connected to the back of
the facing. The advantages of the GRS integral bridge are presented.
It is proposed to apply these new technologies to the recovery ofembankments, RWs and bridges that were damaged by earthquakes,
floodings, storms and tsunami.

Keywords: Bridge abutment, Full-height rigid facing,
Geosynthetic-reinforced soil, Retaining walls, Staged
construction

Fig 1. GRS RW having a FHR facing for Yamanote Line(one of the busiest urban railways in Japan) flying over Chuo
Line (another busiest urban railway) in Tokyo (constructedduring 1995–2000): a) typical cross-section; b) wall underconstruction; and c) completed wall.

1. INTRODUCTION
Construction of geosynthetic-reinforced soil retaining walls(GRS RWs) and geosynthetic-reinforced steep-sloped
embankments has become popular these two decades in Japan.
A couple of unique technologies of GRS structure weredeveloped in Japan, including several new type bridge
abutments comprising geosynthetic-reinforced backfill.

RRR-B:855
15

1.1 GRS RWs having a staged-constructed FHR facing
GRS RWs having a stage-constructed full-height rigid (FHR)
facing is now the standard RW construction technology for
railways including bullet train lines in Japan, replacing
traditional type RWs (Tatsuoka et al., 1997, 2007). Fig. 1
shows a typical wall. This new type GRS RW has been
constructed at more than 850 sites in Japan, and the total wall
length is more than 123 km as of March 2011 (Fig. 2).

4795a) 2041830

154 191

131
Total

Annual wall length （km）

120
100

20
Annual

80
60

10

40

0
1990

1995

2000

2005

2010

Total （km）

This new type GRS RW has the following features (Fig. 3):a) The use of a FHR facing that is cast-in-place using staged
construction procedures. The geosynthetic reinforcement
layers are firmly connected to the back of the facing, whichis essential for high static and dynamic wall stability, as
illustrated in Fig. 4.
b) The use of a polymer geogrid for cohesionless soil toensure good interlocking and a composite of non-woven
and woven geotextiles for high-water content cohesive

140

Fiscal year in Japan (from April to next March)b)
Fig. 2. a) Locations; and b) length of GRS RWs with astaged-constructed FHR facing (as of March 2011).

\

3

facing and the main body of the geosynthetic-reinforced
backfill the following two mechanisms. Firstly, the fresh
concrete can be easily penetrated into the inside ofgravel-filed gabions through openings of the geogrid.
Secondly, extra water from fresh concrete is absorbed bygravel inside the gabions, which prevents negative effects of
bleeding phenomenon of concrete. It is to be noted that thegabions wrapped-around with geosynthetic reinforcement
and filled with gravel function that are placed at the shouldersof soil layers functions as; not only a) a temporary facing
structure during construction that makes backfill-compaction
more easily and resists against earth pressure generated bycompaction and further backfilling at higher levels of thewall; but also b) a drainage layer after construction; and c) a
buffer that protects the connection between the FHR facing
and the reinforcement layers against relative displacementthat takes place after construction. Moreover, concrete formon both sides of the facing and its propping, which becomes
more expensive at a high rate as the wall becomes higher, isnecessary to construct a conventional RC cantilever RW. On
the other hand, only external concrete form without any
external propping while not using internal concrete form isnecessary with this new GRS RW system (Fig. 3b).

c) The use of relatively short reinforcement, made possible byusing planar geosynthetic reinforcement, which has arelatively short anchorage length necessary to activate thetensile rupture strength.
Drain hole

1) Leveling pad for facing

Gravel gabion
Geosynthetic

2) Placing geosynthetic &gravel gabions

Lift = 30 cma)3) Backfilling & compaction

4) Second layer5) Completion of
wrapped-around wall6) Casting-in-place
RC facing

Unstable active zone

Very stable active zone

Connectedb)

Reinforcementc)

Low confiningpressure

Low tensile force

Steel reinforcement

No connection strength

High confiningpressureHigh tensile force

High connection strength

Fig. 4. Effects of firm connection between the reinforcement
and the facing (Tatsuoka, 1992).

Head of anchor steel rod

Fig 3. Staged construction of a GRS RW: a) respective
construction steps; b) details of connection between thefacing and the reinforced backfill; and c) typical view ofthe wall face before casting-in-place FHR facing.

The importance of connecting reinforcement layers to therigid facing is illustrated in Fig. 4. If the wall face is loosely
wrapped-around with geosynthetic reinforcement without
gabions placed at the shoulder of respective soil layers or thereinforcement layers are not connected to a rigid facing, no
tensile force is activated at the wall face in the reinforcement
while no significant earth pressure is activated at the wall face.
No earth pressure at the wall face means no lateral confiningpressure activated to the active zone of the backfill, which
results in low stiffness and low strength of the active zone andtherefore intolerably large deformation and displacement of
the active zone (Fig. 4a). On the other hand, with this new
GRS RW system, as the gabions function as a temporary
facing structure, high earth pressure can be activated at thewall face before placing a FHR facing. As wrapping-aroundgeosynthetic reinforcement at the wall face is buried in the
fresh concrete layer, eventually the reinforcement layers are
firmly connected to the FHR facing and the earth pressurethat has been activated to the temporary facing structure

The staged construction consists of the following steps (Fig.
3a): 1) a small foundation element for the facing isconstructed; 2) a full-height GRS wall with wrapped-aroundwall face is constructed by placing gravel-filled bags at theshoulder of each soil layer; and 3) after the major part ofultimate deformation of the backfill and the subsoil layer
beneath the wall has taken place, a thin (i.e., 30 cm or more)
and lightly steel-reinforced concrete facing is constructed bycasting-in-place fresh concrete directly on the wall face,which makes the FHR facing firmly connected to the main
body of the backfill. Fig. 3b shows the details of theconnection of the FHR facing to the main body of thereinforced backfill and Fig. 3c shows typical view of the wall
face before casting-in-place FHR facing at the site presented
in Fig. 1. A good connection can be made between the RC
4comprising gabions wrapped-around with a geogrid isresisted by a FHR facing consisting of a lightly steel-concrete
layer and a pile of gabions. The importance of this firm
connection for a high wall stability is illustrated in Fig. 4b.
That is, relatively large earth pressure, similar to the active
earth pressure develops in the unreinforced backfill retainedby a conventional RW, can be activated on the back of the
FHR facing because of a high connection strength between
the reinforcement and the facing and a high facing rigidity.
This high earth pressure results in high confining pressure in
the backfill, therefore high stiffness and high strength of thebackfill, which results in high performance of the wall.

overturning moment and lateral thrust force develops at thebase of the facing. A large stress concentration may develop
at and immediately behind the toe on the base of the facing,which makes necessary the use of a pile foundation in usualcases. These disadvantages become more serious at a highrate with an increase in the wall height.
Relatively large earth pressure, similar to the one activated on
the traditional type RW, may be activated on the back of the
FHR facing of this new type GRS RW because of firm
connection between the reinforcement and the FHR facing.
Despite the above, as the FHR facing behaves as a continuous
beam supported at many levels with a small span, typically 30
cm, only small force is activated in the facing structure (Fig.
6). Because of the above, the facing structure becomes rather
simple while the overturning moment and lateral thrust force
activated at the facing base becomes small, which makes
unnecessary a pile foundation in usual cases. The casehistories until today have validated that the GRS RW having a
stage-constructed FHR facing is much more cost-effective(i.e., much lower construction cost, much speedy construction
using much lighter construction machines), therefore a muchless total emission of CO2 than the traditional type RW.
Despite the above, the performance of the new type GRS RW
is equivalent to, or even better than, that of traditional type
RW.

The geosynthetic reinforcement that is required to maintainthe stability of GRS RW having a staged constructed FHR
facing becomes relatively short when compared to metal strip
reinforcement. This is because: 1) the anchorage length ofplanar geosynthetic reinforcement to resist against the tensile
load equal to the tensile rupture strength of reinforcement
becomes much shorter; and 2) a FHR facing prevents theoccurrence of local failure in the reinforced zone of thebackfill by not allowing failure planes to pass through thewall face at an intermediate height. Factor 2) becomes more
important when the backfill is subjected to concentrated loadon the top of the facing or immediately behind the wall faceon the crest of the backfill.

Earthpressurea) Cantilever type RW

Large force in the facing &;large overturning moment &large lateral load at thefacing bottom

5. RC wall

6. Backfill

2&3. Excavation

2&3. Anchor*

Needs for a massive &strong facing; & a pilefoundation
Stress concentration

4. Pile1. Sheet pile*Fig. 5. Traditional type RW as a cantilever structure.
* Temporary structures

a1)
Earth
pressure
Reinforcement

Cantilever type RWVery small force
in the facing⇒ simple facing
structure

Concrete form
& propping

Very small overturning moment & lateralforce at the bottom of facing⇒ usually, no need for a pilefoundation!

a2)b) GRS-RW with a stagedconstructed FHR facing

Fig. 6. GRS RW with a FHR facing as a continuous beam
supported at many points with a small span (Tatsuoka,
1992; Tatsuoka et al., 1997).

2. GRS wall
3. FHRfacing
1. Excavationb)
Fig. 7. Reconstruction of a gentle slope of embankment to avertical wall: a1) & a2) the traditional method; and b) the
new method (the numbers indicate construction sequence).

1.2 GRS-RWs as non-cantilever strutures
A conventional type RW is basically a cantilever structurethat resists against the active earth pressure from theunreinforced backfill by the moment and lateral thrust force atits base (Fig. 5). Therefore, large internal moment and shear
force is mobilized in the facing structure while large
5are connected to the facing unit. Moreover, a foundation thatis embedded inside the reinforced backfill can exhibit large
lateral resistance against the lateral load acting in the
direction normal to the wall face (Fig. 9a). The features of
FHR rigid facing described above become most advantageouswhen a FHR facing functions as a foundation for a
super-structure. The most typical application is bridge
abutments made of GRS RWs having a staged constructed
FHR facing, as described in Chapter 4.
a)b)

Two-directional geogrid

c)

V
L

L
FHR facing
Plan

Vertical construction
joint in the RC facinga)d)Fig. 8. a) Before; b) during; and c) after reconstruction ofembankment slope to a GRS RW having a FHR facing; and d)
typical cross-section, a yard for bullet trains (Shinkan-Sen) in
Nagoya; average wall height= 5 m; total length= 930 m; &construction period= 1990 -1991 (Tatsuoka et al., 1997).

b)Fig. 9. 3-D resistance of FHR facing/geogrid system against
lateral load acting to; a) a vertically long structure locatedinside the reinforced zone; and b) the top of the facing(Tatsuoka et al., 1997).
1.4 Summary
No case of collapse and excessive deformation has beenreported among many case histories of this new type GRS
RW (Fig. 2). This may be attributed mainly to the following
factors: a) a good compaction of the backfill is ensured due toa relative small vertical spacing of geogrid layers (i.e., 30 cm)
and no rigid facing existing during backfill compaction; b) all
potential problems due to deformation and displacements ofwall and ground can be recognised and dealt with before theconstruction of a FHR facing; c) gabion bags stacked
immediately behind the FHR facing ensure good drainage
and act as a buffer when relative displacement tends to takeplace between the facing and the reinforced backfill after
having been opened to service; d) the use of planar geogrid,
rather than metal strips (which are much easier to pull out);
and f) the GRS RWs are designed against high seismic loads;
it was confirmed that duly designed and constructed GRS
RWs can survive such severe earthquakes as the 1995 Kobe
Earthquake (Tatsuoka et al., 1998), as described in Chapter 3.
The design rupture strength of geogrid is usually determinedby the a-seismic design, and, therefore, the design rupture
strength is not reduced to account for creep rupture bylong-term static loads. Yet, no case history in which the wall
has exhibited noticeable creep deformation has been reported.

1.3 Reconstruction of slopes of existing embankments
One of the reasons for a popularity of this new type GRS RW
is a high cost-effectiveness when reconstructing gentle slopesof existing embankment to vertical RWs, compared with thetraditional method (Fig. 7a). In particular, when the stiff
bearing soil layer is deep, expensive temporary structures (i.e.,
ground anchor, sheet piles and concrete form with its
propping) becomes necessary with the traditional method. On
the other hand, with the new method (Fig. 7b), such
temporary structures as listed above are not used while thenumber of construction steps is much smaller, the occupied
space is much smaller and the construction period is much
shorter. Fig. 8 shows a typical case history. Moreover, by the
new method, taking advantage of a FHR facing supported byreinforcement layers for a full wall height, super-structuresthat may exert large lateral load, such as electric poles andhigh noise barrier walls, can be constructed either
immediately behind the wall face without a deep pilefoundation (Fig. 8d), or directly on the FHR facing. In thisrespect, three-dimensional effects of FHR facing (Fig. 9,
make a GRS RW very strong against concentrated load
applied to the top of the facing. That is, the FHR rigid facingof this new GRS RW system is continuous not only in thevertical direction but also in the lateral direction. One unit of
FHR facing, separated from horizontally adjacent units byvertical construction joins in the facing concrete, has somelength, about 10 m. Therefore, the whole FHR facing unit can
resist against concentrated vertical or lateral load applied to
the facing top with a help of all the reinforcement layers that

2. RECONSTRUCTION OF EMBNAKMENTS ANDRETAINNG WALLS COLLAPSED BY NATURALDISASTERS
2.1 Collapse by earthquakes
Numerous embankments and traditional type RWs collapsedby floodings, earthquakes and storms in the past in Japan (e.g.,6Fig. 10). Previously, most of the collapsed soil structures
were reconstructed to respective original traditional types
despite that they are not cost-effective and their resistanceagainst natural disasters is insufficient. From the beginning of
the 1990’s, reconstruction of railway embankments thatcollapsed by heavy rainfalls and floodings to embankments
having geosynthetic-reinforced steep slopes or GRS RWs
having a stage-constructed FHR facing or their combination
started based on the successful experiences described above.

that totally collapsed during the 2004 Niigata-ken Chuetsu
Earthquake. GRS RWs having a FHR facing wereconstructed at these three sites because of not only much
lower construction cost and much higher stability (inparticular for soil structures on a steep slope), but also much
faster construction and a significant reduction of earthwork
when compared to reconstruction to the original
embankments. The new type GRS RWs are also much morecost-effective and the construction is faster than bridge typestructures. During this earthquake, road embankments
collapsed at numerous places in mountain areas and many ofthem were reconstructed to GRS RWs or embankments
having geosynthetic-reinforced steep slopes.
After remedy work:
GRS-RW with a FHR facing; slope: 1:0.3 (V:H);
height= 13.2 m, vertical spacing of geogrid= 30 cm

Fig. 10. Gravity type RW without a pile foundation at
Ishiyagawa that collapsed during the 1995 Kobe
Earthquake (Tatsuoka et al., 1997, 1998).

Failedgravity RW
Shinano riiver

After
failure13.18 mBefore failure: sand backfill including
round-shaped gravel on sedimentary softrock (weathered, more at shallow places)
After remedy
work

1:2
0

.

Silt rock
Sand rock:H)
1:4 (Va)a)

Railway(Jo-etsu line)

Rock bolt

Gravel-filled steel wire
mesh basket

8 July 1992b)c)Fig. 12. a) Railway embankment that collapsed during the2004 Niigata-ken Chuetsu Earthquake and its
reconstruction to a GRW RW having a FHR facing; b) thewall during reconstruction; and c) the completed wall
(Morishima et al., 2005).

b)

The March 25th 2007 Noto-hanto Earthquake caused severe
damage to embankments of Noto Toll Road (opened in 1978).
The north part of this road runs through a mountainous area
for a length of 27 km. The damage concentrated into this part,
where eleven high embankments filling valleys wereextensively collapsed (Koseki et al., 2008). The collapsed
embankments were basically reconstructed to GRS RWs
while ensuring the drainage of ground and surface water.

24 Jan. 1995

Fig. 11. GRS RW having a FHR facing at Tanata; a)
immediately after construction & a typical cross-section;and b) one week after the 1995 Kobe Earthquake (Tatsuoka
et al., 1997, 1998).
High performance during the 1995 Kobe Earthquake of a
GRS RW of this type that had been constructed at Tanata
validated its high-seismic stability (Fig. 11). Many gentle
slopes of embankment and traditional type RWs thatcollapsed by the 1995 Kobe Earthquake and subsequent
earthquakes were reconstructed to GRS RWs having a
stage-constructed FHR facing (Tatsuoka et al., 1977, 1998).
Fig. 12 shows reconstruction of one of the three railway
embankments supported by gravity type RWs at the slope toe

The 2011 Great East Japan Earthquake was the mostdisastrous earthquake in Japan after the World War II. The
damage was a combination of that from seismic motions andthe accompanying tsunami, and was of much greater scalethan that caused by the previous inland fault earthquakes inJapan. A great number of embankments and retaining wallsthat had not be designed and constructed following thecurrent seismic design standard collapsed. In comparison, a7

number of GRS RWs having staged constructed FHR facing
described in Chapter 2 that had been constructed in affectedareas of this earthquakes performed very well. Several
embankments that collapsed were reconstructed to this typeof GRS-RWs (Fig. 13). The details of these case histories will
be reported in the near feature.

Ra ilway track
1V:

26.5 mSecondary low-stiffness
geogrid for compaction control
:
1V

1. 5

1. 5
H

H

Geogrid (rupture strength
TR = 29 .4 kN/m)

1V:0.2H

7m

Gabions between the facing andthe backfill and a large-diameterdrainage pipe are not shown.

b)

0.65 m 11 m

Geogrid (TR = 58 .8kN/m)a)c)Fig. 14. a) Railway embankment damaged by rainfall in 1989;b) reconstructed cross-section; and c) after reconstruction in1991 (Tatsuoka et al., 1997; 2007).
A great number of embankments for roads and railways
retained by gravity-type or leaning type RWs along rivers andseashores often collapsed during floodings and storms. This isdue usually to over-turning failure of the RWs caused byscouring in the supporting ground (Fig. 15a). This type of
failure results from the fact that the conventional type RW is acantilever structure of which the stability is controlled by thebearing capacity at the bottom of the RW (Fig. 5). On theother hand, GRS-RWs with a FHR facing is not such acantilever structure as above (Fig. 6), therefore, much morestable and cost-effective (Fig. 15b). Fig. 16a this type of
failure of a gravity-type RW for a length of 1.5 m along a sea
shore facing the Pacific Ocean, west-south of Tokyo. The
failure was triggered by scouring in the supporting ground bystrong wave actions during a typhoon September 2007. The
wall was restored by the GRS-RW technology. The FHR
facing has a strong resistance against storm waves and stableagainst scouring in the supporting ground. Reinforced soil
RWs with a discrete panel facing is not relevant in such cases,
as the lost of stability of a single panel by, such as erosion of
the backfill from connections between adjacent panels, mayeasily result into a failure of the whole wall.

b)
Fig. 13. a) Embankment collapsed in the Nagano-Niigata
Border Earthquake induced on the day after this disaster; andb) reconstruction to embankment retained by a GRS-RW
having a FHR facing, between Yokokura and Morinomiya,
JR East Iiyama Line (by the courtesy of the East Japan
Railway Co.)
2.2 Floodings and storms
The GRS-RW technology described in Chapter 2 has also
been used to restore soil structures that collapsed by floodings
and storms. Fig. 14 shows a typical case. Railway
embankments that had been constructed in several narrowvalleys collapsed by over-flow of flood water of heavy
rainfall. These embankments were reconstructed by using the
GRS-RW technology. A large diameter drainage pipe wasarranged crossing the respective embankments. Theinteraction between the drainage pipe and the FHR facing wasvery small, because the FHR facing was constructed after
major deformation of the embankment and supporting groundhad taken place. This restoration method was employed also
in many other similar cases (Tatsuoka et al., 1997).

GRS-RWs with a FHR facing has ahigh resistance against scouring

2. Over-turning of RW
3. Collapse ofembankment
Flood

Flood

River bed
1. Scouring

Scouringa)b)Fig. 15. Schematic diagrams showing: a) collapse of
conventional type RW by flooding (the numbers show thesequence of events); and b) high performance of GRS-RW
with a FHR facing

a)

8

b)

a)
c)

b)

d)
Fig. 17. Restoration of the earth-fill dam of Hirata Ike,
Ishikawa Prefecture, that collapsed by the 2007 Noto-Hanto
Earthquake (Mohri et al., 2009).
Relocation of residential areas e to higher ground

Residential areas

Industrial areas

Road

c)

Evacuation buildings(for industrial use)
Coastal dike

Nailed slope withgood drainage
Original
ground surface

Residential areas

Sea
Geosynthetic-reinforced
backfill with good compaction
and drainage

Multiple Tsunami Defense Facilities
Railway

FarmlandGreen belt

Road
Coastal diked)Fig. 16. Seawall of Seisho by-pass for National Road No. 1 in
Kanagawa Prefecture, southwest of Tokyo: a) collapse for alength of about 1 km by Typhoon No. 9, 29th Aug. 2007; b)
original structure; c) reconstructed seawall; and d) the wall
during construction (a, b & c: by the courtesy of the Ministryof Land, Infrastructure, Transport and Tourism)

Nailed slope withgood drainage

Sea
Original groundsurface

Geosynthetic-reinforced
backfill

Fig. 18. Multiple tsunami defense facilities and relocating
residential areas to higher ground using soil-reinforcement
technology
3.3 Possible applications to restoration from the damage
by the 2011 Great East Japan Earthquake
To counter the danger of tsunami, multiple tsunami defense
systems could be considered as well as relocating residential
areas to higher ground. In this earthquake, a great number ofcoastal dikes covered with concrete slabs and panels were
washed away because of overflow of deep tsunami. To
prevent such overflow, embankments that are high enoughmight be constructed with a gentle slope using conventional
technology. Then the width of the embankments and thequantity of earthworks would be extremely large. Where roadand railway embankments are expected to function as
secondary tsunami barrier and evacuation location, a specificheight will be required. Again, if normal embankments withgentle slopes are constructed, they will be wide and involve
large quantities of earthworks. Another proposal is to relocate

Every year, a number of old small scale irrigation damscollapsed by earthquakes and floodings in Japan. Fig. 17
shows the restoration of such a dam as above that collapsedby the 2007 Noto-Hanto Earthquake. The spillway section
was reconstructed by reinforcing the fill by using soil bags
having a tail and a wing to be integrated with the adjacent soil
bags when stacked to have a high resistance against over-flow
and seismic load.

a)9residential areas to higher ground. However, in thisearthquake, a great number of embankments and retaining
walls in residential areas that had been constructed using oldtechnology collapsed in and around Sendai City.

bearings) and connections between simple-supported girders
are generally costly. Moreover, the bearings and connections
are weak points when subjected to seismic loads. Fourthly, abump may be formed behind the abutment by long-term
settlement of the backfill due to its self weight and trafficloads. Lastly, the seismic stability of the backfill is relatively
low; the backfill may deform largely by seismic loads.
Furthermore, the abutment supporting the girder via a fixedshoe is also relatively low, while the girder may dislodge at a
movable shoe.

To make the projects of multiple tsunami defense facilities
and relocating residential areas to higher ground successful, it
is proposed to use the technology of GRS-RW with a FHR
facing explained in Chapter 2, in addition to the execution of
appropriate compaction control and provision of suitabledrainage. Figs. 18a and b schematically show this proposal.
4 DEVELOPMENT OF GRS INTEGRAL BRIDGE

Traditional type

4.1 Problems with conventional type bridges
A traditional type bridge comprises a single simple-supported
girder supported by a pair of abutments via fixed (or hinged)
and moveable bearings (or bearings), or multiple
simple-supported girders supported by a pair of abutments
and a single or multiple pier(s) via bearings. The traditional
type abutment, which may be a gravity structure(unreinforced concrete or masonry) or a RC structure, has thefollowing many drawbacks (Fig. 19).

IntegralGRS RW
Combined

GRS Integral(7) Long-term service: 7a. settlement by self-weight & traffic load; &7b. large deformation by seismic load

Fig. 20. Development of new type bridges.

(6) Girder (deck)(5) Shoes (or bearing)(fixed or movable);
5a. Long-term maintenance5b. Low seismic stability

To solve several problems
with RC & steel structures

To solve several
problems with backfill(3) Backfill(2)
abutment(7) Long-term service: a. settlement by self-weight & traffic load; &b. large deformation by seismic load(4a) Earth pressure(static & dynamic)(4a) Displacement due toearth pressure(3) Girder(4) Integration

Supporting
groundNumbers (1) – (7):event sequence.
(1) Piles(5) Backfill(2) RC abutment(4b) Ground settlement andlateral flow due to the weightof backfill, and associatednegative effects on the piles(i.e., negative friction &bending).

(6a) Displacement due
to earth pressure

Fig. 19. Several major problems with conventional typebridge abutments.

(1) Pilesa)Firstly, as the abutment is a cantilever structure that retains
unreinforced backfill (Fig. 5), earth pressure activated on its
back induces large internal force as well as large thrust force
and overturning moment at the bottom of the abutment.
Therefore, usually, the abutment becomes massive, while apile foundation becomes necessary unless the supporting
ground is strong enough. This drawback becomes more
serious at an increasing rate with an increase in the wallheight. Secondly, although only small movement is allowed
once constructed, the backfill are constructed after theabutments are constructed. Hence, when constructed on soft
ground, many piles may become necessary to prevent
movements of the abutments due to earth pressure as well as
settlement and lateral flow in the subsoil caused by the
backfill weight. Large negative friction may develop along
the piles. The piles may become much longer than the wallheight when the soft ground is thick. Thirdly, the construction
and long-term maintenance of girder bearings (i.e., girder(6a) Earth pressure(static & dynamic)(6b) Ground settlement and lateral flow due to theweight of backfill, and associated negative effects(i.e., negative friction & bending) on the piles.

T: Seasonal expansion andcontraction by thermal effects

Settlement
due to T

Increase in the earth pressure
due to T; and associatedstructural damage to the facing
and pushing out of the facingb)
Fig. 21. Integral bridge: a) construction sequence andassociated problems; and b) a new problem.
4.2 Integral bridge and GRS-RW bridge
To alleviate these problems with the conventional bridge type,three new bridge types have been proposed (Fig. 20). The
10the reinforcement layers connected to its back is small due totheir shallow depths.

integral bridge has been proposed to alleviate problems withthe structural part (usually reinforced concrete) of the
traditional type bridge. This type is now widely used in theUK, the USA and Canada, mainly due to low construction
and maintenance cost resulting from no use of bearings andthe use of a continuous girder. Furthermore, the seismic
stability of integrated bridge is higher than the traditional type(as shown later). However, as the backfill is not reinforced,
thus not unified to the structural part, the backfill and thestructural part do not help each other. Therefore, this new
bridge type cannot alleviate some old problems with thetraditional type bridges (Fig. 21a) and their static and seismic
stability is not very high. Moreover, as the girder is integrated
to the abutments, seasonal thermal expansion and contractionof the girder results into cyclic lateral displacements of theabutments (Fig. 21b). This results in; 1) development of highearth pressure on the back of the abutment (i.e., facing); and2) large settlements in the backfill (England et al., 2000). The
effects of daily thermal effects are negligible.
3

5. Girder (or deck)a)

3: Sill beam

4. Movable and fixedgirder shoes (orbearings)
1b)
1: GRS RW

2: FHR facing

１： Soil Backfill
１.
５

1- The numbers indicate construction sequence.
- Not to scalea)

Cement-mixed gravel

5. Girder (deck)

1

１：

2

Soil Backfill
１.
５

Cement-mixed gravelc)Fig. 23. A bridge abutment at Takada, Kyushu, for a newbullet train line (Tatsuoka et al., 2005); a) structural details;b) lateral and vertical loading tests of the facing; and c)staged construction .

3: Sill beam: 3a. Relatively low seismic stability
3b: Long-term settlement due tolow vertical stiffness of the backfill

4: Movable and fixed girdershoes (or bearings)
4a. Long-term maintenance4b. Low seismic stability

3

1: GRS RW

2: FHR facings(abutments)b)
Fig. 22. GRS-RW bridge: a) construction sequence; and b)
unsolved old problems.

Taking advantages of the stage-construction procedure (Fig.
3), many bridges comprising a pair of GRS RWs with a FHR
facing that support a simple-supported girder wereconstructed (Tatsuoka et al., 1997). Although this bridge type,called the GRS-RW bridge (Fig. 22), is more cost-effective
than the traditional type (Fig. 19), it has the following
problems: 1) The length of the girder is limited due to low
stiffness of the backfill supporting the sill beam. 2) The
construction and long-term maintenance of bearings is costly.
Moreover, the bearings are weak against seismic loads. 3)
Although the seismic stability of GRS RWs with a FHR
facing is very high (e.g., Tatsuoka et al., 1998; Koseki et al.,2006), it is not the case with the sill beam supporting thegirder via a fixed shoe, because the mass of the sill beam ismuch smaller than the girder while the anchorage capacity of
11

It was then proposed and adopted to place a girder on the topof the FHR facing via a shoe (Watanabe et al., 2002; Tatsuoka
et al., 2005). This proposal was adopted by Japanese railway
engineers and the first prototype was constructed for a newbullet train line in Kyushu (Fig. 23). The abutment wasconstructed by the staged construction procedure placing agirder on the top of the RC facing via a fixed shoe. The
conventional type RC abutment (Fig. 19) laterally supports
the unreinforced backfill, which may activate large static anddynamic earth pressure on the back of the facing. In contrast,with this new type abutment, the reinforced backfill laterally
supports a thin RC facing that supports the girder, without the
backfill activating static and dynamic earth pressure on thefacing. Following this project, a number of similar bridge
abutments (nearly 60) were constructed recently and arebeing planned. Despite the above, this type of abutment is notfree from several problems due to the use of girder bearings.
4.3 GRS integral bridge
To alleviate these many problems with the traditional typebridge (Fig. 19) as well as those with the integral bridge (Fig.
21) and the GRS-RW bridge, Fig. 23) and its improved

version (Fig. 23), Tatsuoka et al. (2008a & b, 2009) proposed
another new type bridge, called the GRS integral bridge (Fig.
24). This new type bridge combines the integral bridge and
the GRS-RW bridge taking advantages of their superiorfeatures while alleviating their drawbacks.

tensile force can be mobilized in the reinforcement during theconstruction of GRS walls (w/o a FHR facing).
Bridge type

Cost & period Maintenanceof construction cost

１

３
３
GRS Integral

a)

Total

１ F, G３

252 gal*２F

5641 gal*

1 C, D’ ２ G’

6

３

９

589 gal*GRS RW
2. PilesD, E

B

Integral

3: FHR facing(abutment)

C, D

１

2

1: GRS RW

1

１
A, B

Traditional

Seismic
stability３
1,048 gal*(* Failure acceleration in model shaking table tests)

A= needs for massive abutments because of cantilever structure.
B= needs for piles because of; 1) cantilever-structure abutments; and2) limited allowable post-construction displacements despite theconstruction of backfill after piles & abutments.
C= high cost for construction and long-term maintenance of girder
bearings (i.e., bearings) and their low seismic stability.
D= long-term backfill settlement by self-weight and traffic loads.
D’= long-term settlement of sill beam.
E= lateral cyclic displacements of the abutment caused by thermal
expansion and contraction of the girder, resulting in high earthpressure and large backfill settlement by the dual ratchetmechanism.
F= large backfill settlement and large dynamic earth pressure
G= low seismic stability due to independent performance of two
abutments
G’= low seismic stability of the sill beam.

4. Continuous girder
5. Integration
1: GRS RW

1

3: FHR facing(abutment)

2. Piles- The numbers indicate construction sequence.
- Not to scale (note: girders may be much
longer than the wall height in actual cases)b)
Fig. 24. GRS integral bridge.
The GRS integral bridge has the following features:
1) The backfill is reinforced with geosynthetic
reinforcement layers that are firmly connected to theback of the FHR facings (i.e., the abutments). If it isnecessary to ensure very high performance of the bridge,
particularly during severe earthquakes, the backfill may
be cement-mixed.
2) The abutments are constructed by the following staged
construction procedure:a) A pair of GRS walls (without a FHR facing, the wall facebeing wrapped-around with geogrid reinforcement) are
first constructed.
b) Pile foundations to support the FHR facings areconstructed, if necessary. If the deformation of the
supporting ground by the construction of the backfill is
not significant, pile foundations may be constructed
before the GRS walls for better constructability.
c) FHR facings are constructed by casting-in-place fresh
concrete on the wall face.
d) A continuous girder is placed on and integrated to thecrest of the facings.
This staged construction procedure is a modification of the
one described in Fig. 3 and, therefore, it has the sameadvantages. That is, firstly the connection between thereinforcement and the facing is not damaged by differential
settlement between the facing and the backfill during wallconstruction. Then, construction of abutments on relatively
compressible subsoil without using heavy piles becomes
possible. Secondly, by compacting well the backfill allowing
sufficient outward movements at the wall face, sufficientFig. 25. Rating of four different bridge types based on cost &performance (the higher points, the higher rating).
With conventional type bridges (Fig. 19) and GRS-RW
bridges (Fig. 22), the length of a single simple-supported
girder is restricted to avoid excessive lateral seismic load tobe activated to the abutment on which a fixed bearingsupports
the girder. With integral bridges (Fig. 21), the girder length islimited to avoid excessive large cyclic lateral displacements
at the top of the abutments by seasonal thermal expansion andcontraction of the girder. The girder length is restricted also tolimit the lateral seismic load activated to the abutments. With
the GRS integral bridge, such restrictions as above are muchlooser, therefore, the actual length of the girder relative to theabutment height could be generally much longer than the one
depicted in Fig. 24. The girder length limit for GRS integral
bridges would be larger than the value for conventional type
integral bridges, which is presently specified to be 50 - 60 m
in the USA to restrict the maximum thermal deformation of
the girder to four inches (about 10 cm). More research will benecessary in this respect.
Many series of model tests (cyclic static loading tests andshaking table tests) were performed to validate theadvantageous feature of the GRS integral bridge over othertypes of bridges, as reported in details by Tatsuoka et al.
(2009). Fig. 25 compares the characteristic features of the
12four different bridge types described in the precedent sections.
The rating presented in this figure is only an approximation.
That is, the full point allocated to each item is three, which is
reduced one by one when any of the listed negative factors A G is relevant. The horizontal accelerations at which therespective bridge models collapsed in the shaking table tests
explained below are listed. A total full point equal to nine isgiven only to the GRS integral bridge.

a)

a)

b)
Fig. 27 The first GRS integral bridge for the new bullet train
line at Kikonai between Shin-Aomori and Shin-Hakodate
stations: a) overall dimensions; and b) a picture at the end of
August 2011 (by the courtesy of Mr. Yamada, Y.).
b)

Fig. 28 A view from the upstream of Tsuyagawa Bridge,
between Motoyoshi and Rikuzenkoizumi stations, Kesenuma
line, East Japan Raiway (taken by the author)c)Fig. 26 A full-scale model of GRS integral bridge underconstruction (Nov. 2008) at Railway Technical ResearchInstitute, Japan: a) dimensions; b) abutment underconstruction; and c) completed model.
A full-scale model of GRS integral bridge was constructed atRailway Technical Research Institute in 2008 (Fig. 26). The
model was constructed taking advantage of a pair of full-scalemodels of GRS-RW with a FHR facing constructed 1998. A
high constructivity of GRS integral bridge was confirmed.
The long-term behaviour is now being observed. In 2011, thefirst prototype GRS integral bridge is now under construction
for a new bullet train line at Kikonai between Shin-Aomori
and Shin-Hakodate stations, at the south end of Hokkaido(Fig. 27). The bridge is heavily instrumented to observe thelong-term behaviour. The details of the construction andbehaviour during and after construction will be repored in thenear future. Although the span is not long, it is considered that
this case history is the historical first step for a number of
GRS integral bridges to be constructed in the coming years.

Fig. 29 A view from the seaside of Namiitagawa bridge,
between Namitagawa and Kirikiri stations, Otsuchicho,
Yamada Line, East Japan Raiway (taken by the author).
A great number of bridges lost their girders by strong tsunami
forces during the 2011 East Japan great earthquake (Figs. 28& 29). These girders, for which measures had been taken toprevent dislodging from the abutments but not to resist
against the tsunami forces, were washed away. On the other
hand, most of the short-span single-girders for local roads
located near the modern bridges that lost girders survived
tsunami forces (Fig,. 29). This is likely because theseshort-span bridges are plugged into the RC RWs on both sides
of a river without using bearings: i.e., these bridges are
actually integral bridges. This fact indicates that GRS integral
13bridges may have a sufficient resistance against tsunami
forces. Moreover, as seen from Fig. 29, the unreinforced
backfill of many bridges was washed out by tsunami forces.
On the other hand, the backfill of GRS integral bridges also
has a high resistance against tsunami forces by being
reinforced with geogrid layers connected to the facings. It isstrongly recommended to reconstruct these bridges damagedby tsunami to GRS integral bridges. Furthermore, GRS
integral bridges are particularly relevant for newly
constructed bridges if tsunami forces are to be taken into
account.

6 ACKNOWLEDGEMENTS
This paper is based on a long-term research for about 30 years
performed by the author and his colleagues. The author
sincerely appreciates all helps and efforts of his previous andcurrent colleagues at University of Tokyo and Tokyo
University of Science in performing this long-term research.
The support of Dr Tateyama, M. and his colleagues at the
Railway Technical Research Institute, Japan, is highly
appreciated.

7[1]

REFERENCES

England, G, L., Neil, C, M. and Bush, D, I., Integral Bridges, A
fundamental approach to the time-temperature loading problem,
Thomas Telford, 2000.
[2] Koseki, J., Bathurst, R.J., Guler, E., Kuwano, J. and Maugeri, M.,
Seismic Stability of Reinforced Soil Walls, Proc. 8th International
Conference on Geosynthetics, Yokohama, 2006, Vol.1, pp. 51-77.
[3] Koseki, J., Tateyama, M., Watanabe, K. and Nakajima, S. 2008.
Stability of earth structures against high seismic loads, Keynote Lecture,
Proc. 13th ARC on SMGE, Kolkata, Vol. II.
[4] Morishima, H., Saruya, K. and Aizawa, F., Damage to soils structuresof railway and their reconstruction, Special Issue on Lessons from the2004 Niigata-ken Chu-Etsu Earthquake and Reconstruction,
Foundation Engineering and Equipment (Kiso-ko), October, 2005,
pp.78-83 (in Japanese).
[5] Mohri, Y., Matsushima, K., Yamazaki, S., Lohani, T. N., Tatsuoka, F.
and Tanaka, T., “New direction of earth reinforcement - Disasterprevention for earth fill dam –“, Gesynthtetics International, IS Kyushu
2007 Special Issue, 16(4), 2009, pp. 246-273.
[6] Tatsuoka, F., Roles of facing rigidity in soil reinforcing, Keynote
Lecture, Proc. Earth Reinforcement Practice, IS-Kyushu ‘92 (Ochiai et
al. eds.), 1992, Vol.2, pp.831-870.
[7] Tatsuoka, F., Tateyama, M, Uchimura, T. and Koseki, J.,
Geosynthetic-reinforced soil retaining walls as important permanent
Structures,” Geosynthetic International, 1997, Vol.4, No.2, pp.81-136.
[8] Tatsuoka, F., Koseki, J., Tateyama, M., Munaf, Y. and Horii, N.,
Seismic stability against high seismic loads of geosynthetic-reinforced
soil retaining structures, Keynote Lecture, Proc. 6th Int. Conf. on
Geosynthetics, Atlanta, 1998, Vo.1, pp.103-142.
[9] Tatsuoka, F., Tateyama, M., Aoki, H. and Watanabe, K., Bridge
abutment made of cement-mixed gravel backfill, Ground Improvement,
Case Histories, Elsevier Geo-Engineering Book Series, Vol. 3(Indradratna & Chu eds.), 2005, pp.829-873.
[10] Tatsuoka, F., Tateyama, M., Mohri, Y. and Matsushima, K., Remedial
treatment of soil structures using geosynthetic-reinforcing technology,
Geotextiles and Geomembranes, 2007, Vol.25, Nos. 4 & 5, pp.204-220.
[11] Tatsuoka, F., Hirakawa, D., Nojiri, M., Aizawa, H., Tateyama, M. and
Watanabe, K., Integral bridge with geosynthetic-reinforced backfill,
Proc. First Pan American Geosynthetics Conference & Exhibition,
Cancun, Mexico, 2008a, pp.1199-1208.
[12] Tatsuoka, F., Hirakawa, D., Aizawa, H. Nishikiori, H. Soma, R and
Sonoda, Y., Importance of strong connection between geosynthetic
reinforcement and facing for GRS integral bridge, Proc. 4th
GeoSyntheticsAsia, Shanghai, 2008b.
[13] Tatsuoka, F., Hirakawa, D., Nojiri, M., Aizawa, H., Nishikiori, H.,
Soma, R., Tateyama, M. and Watanabe, K., A new type integral bridge
comprising geosynthetic-reinforced soil walls, Gesynthtetics
International, IS Kyushu 2007 Special Issue, 2009, Vol.16, No.4,
pp.301-326.
[14] Watanabe, K., Tateyama, M., Yonezawa, T., Aoki, H., Tatsuoka, F. andKoseki, J., Shaking table tests on a new type bridge abutment withgeogrid-reinforced cement treated backfill,” Proc. 7th Int. Conf. on
Geosynthetics, Nice, 2002, Vol.1, pp.119-122.

5 CONCLUSIONS
Geosynthetic-reinforced soil retaining walls (GRS RWs)
having a stage-constructed full-height rigid (FHR) facing
have been constructed as important permanent RWs for a
total length of more than 125 km in Japan since 1999 until
today (March 2011). Although these GRS RWs are mainly
for railways, many others were also constructed for highways
and other types of infrastructure. Its current popular use is due
to not only high cost-effectiveness, but also high performancethat is equivalent to, or even better than, other modern RWs.
This success can be attributed to;1) the use of a proper type of reinforcement (i.e., a geogrid forcohesionless soil and a nonwoven/woven geotextile
composite for high-water content cohesive soil);2) casting-in-place of a FHR facing by staged construction
procedures in such that reinforcement layers are firmly
connected to the facing; and3) taking advantages of the rigidity of the facing in design.
Many embankments and traditional type RWs that collapsed
during recent severe earthquakes, heavy rainfalls andassociated floodings and storms, in Japan were reconstructed
to embankments with steep geosynthetic-reinforced slopes or
GRS RWs with a stage-constructed FHR facing or their
combination. It was validated that this technology is highly
cost-effective in restoring collapsed old soil structures.
A new bridge type, called the GRS integral bridge, is
proposed, which comprises an integral bridge andgeosynthetic-reinforced backfill. GRS integral bridges
exhibit essentially zero settlement in the backfill and no
structural damage to the facing when subjected to lateralcyclic displacements of the facing caused by seasonal thermal
expansion and contraction of the girder, while their seismic
stability is very high. These features and highcost-effectiveness of the GSR integral bridge are due to that:
1) bearings are not used; 2) the girder is continuous without
any connections; 3) the backfill is reinforced with geogrid
layers firmly connected to the facing; and 4) FHR facings arestage-constructed after the construction of full-heightgeosynthetic-reinforced soil walls and then pile foundations(if necessary).
14

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Seepage-deformation Coupled Numerical Analysis of Unsaturated River
Embankment Using an Elasto-plastic Model
F.Oka, and S.Kimoto, Department of Civil and Earth Resources Engineering, Kyoto University
R.Kato, Nikken Sekkei Civil Engineering Ltd.


 ijf   nS r p f  ij

ABSTRACT: In the present paper, a water-soil coupledelasto-plastic finite element analysis method is proposed forunsaturated river embankments by incorporating unsaturated
soil-water characteristics. The simplified three-phase method havebeen developed based on the multi-phase porous theory by assumingthat the compressibility of pore air is very high. For the constitutive
model for soils, we have used an elasto-plastic model consideringboth the strain-induced degradation and the change of suction. Using
the proposed method, we have analyzed a model test of the seepage
flow through the unsaturated embankment and compared with theexperimental results. And we have found the proposed method isapplicable for the numerical modeling of unsaturated embankment
with seepage flow. In addition, we have studied the effect of material
parameters on the behavior of unsaturated embankment during theseepage flow by the parametric study. From the parametric study,
the effects of the strain-induced degradation and of the change ofsaturation are clarified.

(2)  n(1  Sr ) p  ij
a
ij

(3)

a

   'ij  (1  n) S r p  ij  (1  n)(1  S r ) p  ij
s
ij

f

a

(4)

where  ij is the skeleton stress [1], p f and p a are the porewater pressure and the pore gas pressure, respectively, n is
the porosity, and S r is the degree of saturation.
The skeleton stress is used as the basic stress variable in themodel for unsaturated soil. Summing up Eqs.(2)-(4) we havethe total stress tensor as(5)
 'ij   ij  P F  ij ,P F  S r p f  (1  S r ) p a
where P F is the average pore pressure.
Adopting the skeleton stress provides a natural application of
the mixture theory to unsaturated soil. Equation (5) is similar
to Bishop’s definition for the effective stress of unsaturated
soil. In addition to Equation (5), the effect of suction on theconstitutive model should always be taken into account. This
assumption leads to a reasonable consideration of the collapse
behavior of unsaturated soil, which has been known as abehavior that cannot be described by Bishop’s definition for
the effective stress of unsaturated soil. Introducing suction( p c  ( p a  p f ) ) into the model, however, makes it
possible to formulate a model for unsaturated soil, startingfrom a model for saturated soil, by using the skeleton stressinstead of the effective stress. An elasto-plastic model [2] hasbeen be extended to unsaturated soil using the skeleton stressas well as the effect of suction.

Keywords: unsaturation, embankment, seepage-deformation
coupled analysis, elasto-plastic model
1. INTRODUCTION
In recent years, river embankments have been failed due toheavy rains in the world. It is well known that the one of thereason for the failure of the river embankment is seepage flow
due to a rise of the river-water table.
In the present paper, we have numerically analyzed thedeformation behavior of river embankment as well as theseepage flow during a rise of the water level using asoil-water coupled analysis method. We have analyzed amodel test of the seepage flow through the unsaturated
embankment and compared with the experimental results. In
addition, we have studied the effect of material parameters on
the behavior of unsaturated embankment during the seepage
flow through the parametric study.

2.2 Mass conservation
The mass conservation law for the three phases is given by J  (  J uiJ )
(6)

0
t
xi
where  J is the average density for the J phase and u iJ isthe velocity vector for the J phase.
(7) s  (1  n )  sff(8)  nS r 

2. SEEPAGE-DEFORMATION COUPLED ANALYSISOF UNSATURATED SOIL
2.1 Stress variables for the mixture
The total stress tensor is assumed to be composed of threepartial stress values for each phase.
(1) ij   ijs   ijf   ija(9) a  n(1  S r )  awhere J =s, f, and g in which super indices s, f, and g indicate

where  ij is the total stress tensor, and  ijs ,  ijf , and  ija

the solid, the liquid, and the gas phases, respectively.  J is
the mass bulk density of the solid, the liquid, and the gas.

are the partial stress tensors for solid, liquid, and gas,
respectively. The partial stress tensors for unsaturated soil canbe given as follows:

2.3 Conservation of linear momentum for the threephases
The conservation law of linear momentum for the threephases is given by15 s uis  Qi  Ri 
 f uif  Ri  Si  auia  Si  Qi  ijsx j
 ijfx j

 ijax j(10)

  f bi(11)

  abi(12)

solid and gas phases, Ri denotes the interaction between
solid and liquid phases, and S i denotes the interaction
between liquid and gas phases.
These interaction terms Qi and Ri can be described as
(13)Ri  nS r wf w if
k
ag
(14)
Qi  n(1  Sr ) a w iakWhen we assume that the interaction between liquid and gas
phases is small and ca be neglected, S i can be described as

Si  0

Incorporating Equation (20) and p f   K f  iif ( K f :volumetric elastic coefficient of liquid,  iif : volumetric strainof liquid) into the above equation leads to the following
continuity equation for the liquid phases:  k f  f s p f
  ui   f bi   S r iis  nS rx i   w 
x i

p f
 nS r f  0 (26)K
Similarly, we can derive the continuity equation as:  k a  a s p a
  a bi   (1  S r )iis  nS r
 a   ui 
xi   g 
xi

p a
 n(1  S r ) a  0 (27)Kais the water permeability coefficient, k is the airpermeability, w if is the average relative velocity vector ofwater with respect to the solid skeleton, and w ia is theaverage relative velocity vector of air to the solid skeleton.
The relative velocity vectors are defined by
f
f
s
(16)
w i  nSr  u i  u i aasw i  n(1  Sr )  u i  u i (17)Using Equation (16), Equation (11) becomes ijf
1 f
  f bi f (uis i )  Ri 
w
x j
nSr

(18)

2.6 Air pressure
When we assume that the air is elastic and that its constitutive
equation is given by p a   K a  iia ( K a : volumetric elastic
coefficient of air,  iia : volumetric strain of air), Equation (27)becomes   k a  a s p a
K a 
  a bi   (1  S r )iis  nSr 
   uii xi
 xi   w a n(1  Sr ) p  0 (28)if  0 and use Equations (3), (6), andWhen we assume that w(16), Equation (18) becomes

p
(19)
nS r  f uis  nS r wf w if   nS r
 nS r  f bikxi
After manipulation, the average relative velocity vector ofwater to the solid skeleton and the average relative velocity
vector of air to the solid skeleton are shown as
k f pf(20)
  f uis   f bi )w i   ( w xiaw i  

k a p(   a uis   abi ) a g xi

When the air compressibility is very large, compared with theother phases, we can set K a  0 . In other words, from
Equation (28), we obtain P a  0 . The above discussion
means that we can assume that P a  0 if initial air pressure
Pinia  0 . This shows that the continuity in Equation (28) isalways satisfied.
Since saturation is a function of the pressure head, the time
rate for saturation is given bydS d d c
1 1 c C c
(29)
nSr  n rp  n Cp pd d dp
n w
w

(21)

2.4 Equation of motion for the whole mixture
Based on the above fundamental conservation laws, we canderive equations of motion for the whole mixture. By adding
Equations (10)-(12) and using Equations (7), (8), and (9)wehaveuis  nS r  f (uif  uis )  n(1  S r )  a (uia  uis )



 ij
x j

 bi

f

2.5 Continuity equations for the fluid phase
Using the mass conservation law for the solid and the liquid
phases, given in Equation (6), and assuming theincompressibility of soil particles, we obtain nSr (u if  u is )
 f
(25)
 Sr  iis  nSr f  nSr  0

xi

(15)

f

is the mass density of the mixture ass       a , and ui is the acceleration vector of thesolid phase.
From the following assumptions,uis ≫ (uif  uis ) , uis ≫ (uia  uis ) ,(23)
the equations of motion for the whole mixture are defined as
 ij
(24)
 uis   bix jswhere bi is a body force, Qi denotes the interaction between

where kwhere

  s bi(22)

16

pore water pressure, is assumed to be constant. Below thewater level, the pore water pressure is given by thehydrostatic pressure. In the transition region between thewater level and the constant suction region, we assume thatthe pore water pressure is linearly interpolated (Figure 1).
When the pressure head is negative, the increase in soil
modulus due to suction is considered.

Vwpc
is the volumetric water content,  
is
V
w
d
is the specific water content.
the pressure head, and C 
dUsing Equation (29), Equation (26) becomes

where  




 nS
C 
  k f  f s p f
  ui   0 (30)  f bi   S r iis   fr 





x i
x i  w 
w 
K


Water pressure＝0

The apparent volumetric elastic coefficient of pore water,
K f , is defined as
1
S
C
(31)
 rf 
fKn w
K
Then, the continuity equation can be written asn  k f  f s p f
  f bi   S r iis  f p f  0
   ui 
xi   w 
xi
K


Suction from initial degree of saturation

Hydrostatic pressure(32)

Fig 1. Initial pressure

3 UNSATURATED SEEPAGE CHARACTERISTICS

4 ELASTO-PLASTIC MODEL CONSIDERING BOTH
STRAIN-INDUCED DEGRADATION AND
CHANGE OF SUCTION

The soil-water characteristic model proposed [3] is used todescribe the unsaturated seepage characteristics for which
effective saturation S e is adopted as   r nS r   r
(33)
Se 

s  r s  r
where  s is the volumetric water content at the saturated

For the constitutive model, we use elasto-plastic model
extended to unsaturated soil [2][5]. In this study, we consider
both the strain-induced degradation and the change ofsuction.

state, which is equal to porosity n , and  r is the residual
volumetric water content retained by the soil at the large
value of suction head which is a disconnected pendular water
meniscus.
In order to determine the soil-water characteristics, effective
saturation S e can be related to negative pressure head through the following relation:
m(34)
Se  1  ( ) n 

4.1 Effect of strain-induced degradation
For considering the strain-induced degradation, we change
overconsolidation boundary surface to Equation (38).

z   dz   d ijP d ijP



1



(39)

2

is the reduction rate of  mbi ),  is material parameter for .
adjust the speed of convergence of  mbi

4.2 Effect of change of suction
For considering the change of suction, we change
overconsolidation boundary surface to Equation (40).

 PC

 1  e 0 P    S IE  S I exp s d  i C  1P  
 




   mbi
 exp
 mb(40)

where S IE is the strength reduction rate of after releasing thesuction, S I is the strength increase rate in initial suctionC   (n  1)( s   r ) S e m (1  S e )(36)
Specific permeability coefficient k r , which is a ratio of thepermeability of unsaturated soil to that of saturated soil, isdefined [4] bykr  Seb 1  (1  Se m )n

1

 (  mbf
where  mbf
 is the convergence value of  mbi  n  mbi
 , n*



1



(38)

*

where  is a scaling parameter which has the dimensions of
the inverse of  , and n  and m determine the shape of thesoil-water characteristic curve. The relation between n  andm leads to an S-shaped type of soil-water characteristic
curve, namely,1(35)m  1
nSpecific water content C , used in Equation (29), can becalculated as1

 1  e0 P v      mbf
   mbi   mbf
  exp z exp
 mb

m m

( S IE  S I  1.0 ), sd is the parameter of adjusting the speed ofchange of strength, Pi C is initial suction, P C is the current
suction. Finally, Overconsolidation boundary surface isdefined as   mbf   mbi   mbf
  exp    z 
 mb

(37)



Applying the above-mentioned relations, we can describe theunsaturated seepage characteristics.
In the analysis, the unsaturated region is treated in thefollowing manner:
In the embankment, the initial suction, i.e., the initial negative




 PC
 
 1  e0 P  
exp v   S IE  S I exp  sd  i C  1  (41)     P 


Next, for considering the change of suction, we change theplastic shear coefficient to function of suction. Then, we
17

define the parameter B* which adjust the speed of kinematic
hardening as(42)

where S IBE is the strength reduction rate of after releasing thesuction, S IB is the strength increase rate in initial suction( SIBE  S IB  1.0 ), sdb is the parameter of adjusting the speed ofchange of strength.

5 NUMERICAL ANALYSIS OF A RIVER
EMBANKMENTIn the numerical analysis we have used a finite element
method to solve governing equations. The computer program
LIQCA2D-SF11 [5] has been developed and used.

5.1 Finite element mesh and boundary conditions
Figure 2 shows the model of the river embankment and thefinite element mesh used in the analysis and the position ofgauges at experiment. The soil parameters for Yodogawa
sand are listed in Table 1, which is a soil used for thereconstruction of levee of Yodo river [6]. For the simplified
model, the effect of suction and the strain-induced
degradation is not considered. Initial saturation ofembankment is 60% and the coefficient of permeability forsaturated soil is shown in Table 2. The air permeability is not
specified because the analysis is simplified such that the airpressure is always zero, as indicated in Equation (28). Taking
account of the previous analysis [7], we set the water
characteristic curve and specific permeability as Figure 3.
The water level-time profile is shown in Figure 4.

Compression index λ0.0804

Swelling index κ
Normalized Ini. shear modulus G0/σ’ｍ０

250.0

2

0.009

Gravitational acceleration g (m/s2)

9.8

Density ρ（t/m3）
Stress ratio at phase transformation Mm＊

1.9
1.270

Stress ratio at failure Mf＊

1.270

Hardening parameter B0＊

800

Hardening parameter B1＊
Hardening parameter Cf

600

Quasi-OCR OCR＊
Anisotoropy parameter Cd

2000

Dilatancy parameters D0* ,n2.0, 2.0

Plastic ref. strain γrefP＊

0.008

Elastic ref. strain γrefE＊

0.08

20
1.3

Surface element on riverside

Embankment
Horizontal

VerticalHorizontal

-5

-6

-7

4.79×10

Displacement gauge

#5
#2

0.535

Table 2.Coefficient of permeability for saturated soil(m/s)

Suction (Head, m)

#4
#1

yodo

Method※Initial void ratio e0#3

8.79×10

8.79×10

-7

1.0

2.0
1.8
1.6
1.4
1.2
1.0
0.8
0.6
0.4
0.2
0.0

0.8

Soil-water characteristic curve
Specific permeability

0.6
0.4
0.2
0.0
0

Water pressure gauge

4.79×10

VerticalSpecific permeability



 PC

*
B suc
 B 0*  S IBE  S IB exp s db  i C  1
P
 



 


Table 1. Material parameters
Soil

0.1

0.2

0.3

0.4 0.5 0.6
Saturation

0.7

0.8

0.9

1

Fig 3. Water characteristic curve and specific permeability.

Water pressure gauge No.

Water level (m)1.50Time (hr)
0

0.13

24

Fig 4. Water level.
Displacement gauge No.

Fig 2. Model of embankment and the finite element mesh.

18

5.2 Results
The distribution of saturation is illustrated in Figures 5. Afterthe water level of the river reached the crest of theembankment, the seepage flow infiltrated from the river side
to the back side. The distribution of the horizontal local
hydraulic gradient after 24 hours is illustrated in Figure 6. The
horizontal local hydraulic gradient at the toe of theembankment on the back side was not over 0.5. In the Guide
for Structural Investigations of River Embankments, theallowable gradient for Piping is 0.5 [8]. And then, theaccumulated plastic deviatoric strain (  p   d p ，

Max 0.43

Fig 6. Distribution of horizontal local hydraulic gradient(After 24 hours).

Max 2.45%

1d p  (deijp deijp ) 2 ) was 2.45% at the toe of the embankment(Figure 7).
The time histories of water pressure both experiment andanalysis at a water pressure gauge point are illustrated in
Figure 8. The level of the pressure rise of the analysis is thesame as that of experiment. After 24 hours, the value of waterpressure of analysis agrees with that of experimental result.
The time histories of displacement of normal to the surface on
the back side of the embankment are illustrated in Figure 9.
Both analysis and experiment, displacement has occurred
rapidly after 5 hours. After 24 hours, the maximum value ofdisplacement of analysis was about 2.4mm and that ofexperiment was 2.8mm. Hence, we have found the proposed
method is applicable for the numerical modeling of
unsaturated embankment with seepage flow.

Pore water pressure
Pw (kPa)

Pore water pressure
Pw (kPa)Fig 7. Distribution of  p (After 24 hours).

After 1 hour

9
8
7
6
5
4
3
2
1
0

9
8
7
6
5
4
3
2
1
0

Experiment

#1
#2
#3
#4
#5

0

5

10
Time (hour)

15

20

Analysis

#1
#2
#3
#4
#5

0

5

10

15

20

25

Time (hour)

Fig 8. Time histories of water pressure.

After 8 hours

Experiment

3
#1
#2
#3
#4
#5
#6
#7
#8
#9
#10

Displacement of
normal direction (mm)2.5After 16 hours

2
1.5
1
0.5
0

After 24 hours

0

5

3

Displacement of
normal direction (mm)

15

20

25

10

15

20

25

経過時間
(h)
Time
(hour)

Analysis
#1
#2
#3
#4
#5
#6
#7
#8
#9
#10

2.5

Fig 5. Distribution of saturation and infiltration line.

10

2
1.5
1
0.5
0
0

5

経過時間
(h)
Time
(hour)

Fig 9. Time histories of displacement of normal direction.
19

Table 3. Analysis cases and the results

6 PARAMETRIC STUDY OF THE EFFECT OF THE
STRAIN-INDUCED DEGRADATION AND OF THE
CHANGE OF SUCTION

SIE

SI

sd

SIBE

SIB

sdB

Strain-induced
After 24hours
degradation
γp
parameter
(％)
*
β
n

－
0.5
－
－
0.5
0.5
－
0.5

－
0.5
－
－
0.5
0.5
－
0.5

－
0.25
－
－
0.25
0.25
－
0.25

－
－
0.5
－
0.5
－
0.5
0.5

－
－
0.5
－
0.5
－
0.5
0.5

－
－
0.25
－
0.25
－
0.25
0.25

－
－
－
0.5
－
0.5
0.5
0.5

Suction parameter

Case

6.1 Finite element mesh and boundary conditions
Figure 10 shows the prototype model of river embankment
and the finite element mesh used in the analysis. The soil
parameters for Yodogawa sand are same as Chapter 4. But,
we set coefficient of permeability for saturated soil as
4.79  105 (m/s) in all elements. The rising rate of the waterlevel is 1/3 (m/hour) until the water level reaches the crest of
the embankment (18 hours), and then it remains constant for
24 hours. And then, we change strain-induced degradation
parameters and/or suction parameters (Equation (41),Equation (42)) in Case1 ~ Case7. The analysis cases and theresults from this study are listed in Table 3. In this table,  p is
the accumulated plastic deviatoric strain at the toe of theembankment on the river side.

0
1
2
3
4
5
6
7

－
－
－
50
－
50
50
50

9.34
11.88
19.03
12.72
20.33
12.30
15.85
15.72

After 9 hours

6.2 Numerical results
The distribution of saturation is illustrated in Figure 11. The
water level of the river reached the crest of the embankment
after 18 hours. Then, the seepage flow infiltrated from theriver side to the land side. The distribution of the horizontal
local hydraulic gradient after 24 hours is illustrated in Figure
12. The horizontal local hydraulic gradient at the toe of theembankment on the land side was more than 0.5.
The distributions of saturation are presented in Figure 13(Case 4, 7). Compared with the result of Case 0, we havefound that strain-induced degradation parameters and thesuction parameters do not affect infiltration. The reason is
because continuity equations are not changed if we consider
both the strain-induced degradation and the change ofsuction.
The distributions of the accumulated plastic deviatoric strain
is illustrated in Figure 14. In the cases, i.e. Cases 2, 4, 6, 7, in
which we take account of the decrease in B* due to thedecrease of suction associated with the water infiltration,
larger strain occurs over the wide area of embankment. The
results listed in Table 3 show that considering thestrain-induced degradation and/or the decrease in suction
leads to the larger value of the accumulated plastic deviatoricstrain at the toe of the embankment on the front side.

After 18 hours

After 24 hours

Fig 11. Distribution of saturation (Case 0).

Max 0.59
20m

12m

5m

12m

Back side
6m
Initial water level

10m

y

20m

Front (river) side
Embankment

Water level(18hours)

Foundation groundFig 12. Distribution of horizontal local hydraulic gradient(Case 0, after 24 hours).

x

Fig 10. Model of embankment and the finite element mesh.

Considering the suction dependency of the parameter B*(Case 2, 4, 6, 7), we can observe comparatively larger strain
20

compared with other cases for the decrease in suction.

7 CONCLUSIONS
We have analyzed a model test of the seepage flow throughthe unsaturated embankment and compared with theexperimental results. From the comparison between
numerical and experimental results, we have found the
proposed method is applicable for the numerical modeling of
unsaturated embankment with seepage flow.
By the parametric study, it has been found that considering
the strain-induced degradation and/or the change of suction
leads to the larger accumulated plastic deviatoric strain at thetoe of the embankment on the front side. In particular, in thecase with the decrease in the parameter B* due to the changeof suction, large strain occurred over the wide area of
embankment. However, strain and/or suction-induced
degradation does not affect infiltration pattern.

Case 4

Case 7

Fig 13. Distribution of saturation (After 24 hours).

Max 9.3%

Max 20.3%

Case 0Case 4

Max 11.9%

Max 12.3%

Case 1Case 5

Max 19.0%

Max 15.9%

Case 2

Case 6

Max 12.7%

Max 15.7%

Case 3

Case 7

Fig 14. Distribution of  p (After 24 hours).
21

8 REFERENCES
[1]

[2]

[3]

[4]
[5]

[6]

[7]

[8]
[9]

Jommi, C., “Remarks on the constitutive modelling of unsaturated
soils”, Experimental Evidence and Theoretical Approaches in
Unsaturated Soils, Tarantino, A. and Mancuso, C. eds., Balkema, 2000,
pp.139-153.
Oka,F, A.Yashima,A.Tateishi,Y.Taguchi and S.Yamashita, “A cyclic
elasto-plastic constitutive model for sand considering a plastic-strain
dependence of the shear modulus”, Geotechnique, 49, 5,1999,
pp.661-680.
van Genuchten,M.T.,”A closed-form equation for predicting thehydraulic conductivity of unsaturated soils”, Soil Sci. Soc. Of Am.
J.,44,1980, pp.892-898.
Brooks, R.H. and Corey, A.T., “Hydraulic properties of porous media,
Hydrology Papers”, Colorado State University, 1964, pp.24.
Kato,R. “Development of unsaturated seeage-deformation coupledsimulation method and its application to river embankment”, PhD
Thesis, Kyoto University, 2011
Kinki Regional Development Bureau, Ydogawa River Office,“ Research report on reinforcement of river embankment against
seepage and overflow”, 2010
Morinaka,Y., “ Multi-phase seepage-deformation coupled analysis ofriver embankment using FEM and MPM ” ,Master Thesis, Graduate
school of Engineering, Kyoto University, 2010.
Japan Institute of Construction Engineering, Guide for Structural
Investigations of River Embankments, 2002.
F. Oka, S. Kimoto, N. Takada, H. Gotoh and Y. Higo, “A
seepage-deformation coupled analysis of an unsaturated riverembankment using a multiphase elasto-viscoplastic theory”, Soils andFoundations, Vol.50, No.4, 2010, pp.483-494.

22

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Progressive Failure of Shallow Anchor in Dense Sand Evaluated by 2D and 3D
Finite Element Analysis and Model Tests
Toshinori SAKAI, Dept. of Environmental Science and Technology, Graduate school of Bioresources,
Mie University, Japan
Md. ROKONUZZAMAN, Dept. of Civil Engineering, Khulna University of Engineering & Technology,
Bangladeshprogressive failure and crushing of sand particles subjected toa high confining pressure. Since anchor uplifting is generally
performed at low confining pressures in conventional 1 g tests,
crushing of sand particles is expected to be negligible during
such tests. The failure of sand mass is usually progressive andit is linked to the development of a narrow shear band of
localized deformation. The appearance of a shear band isfrequently accompanied by a softening phenomenon. Few
studies have employed numerical models that fully consider
progressive failure with shear banding [8,9]. Anchor
problems typically consider three kinds of anchors: strip,
circular, and rectangular anchors, which respectively
correspond to planestrain,axisymmetric, andthree-dimensional problems. Most studies adopt a
two-dimensional design approach for calculating the anchor
uplift resistance [10–12]. Ovesen [13] found that an isolated
anchor did not correspond to a two-dimensional problem.
Meyerhof and Adams introduced an empirical shape factor to
calculate the peak uplift load based on plane strain
assumptions for circular and rectangular anchors. Dickin [14]proposed embedment and shape factors. Isolated anchors are
currently designed based on empirical methods. Few
researchers have proposed rigorous numerical models thatconsider the shape effect. The present study seeks to explain
the progressive failure of a shallow anchor in granularmaterials at the 1 g level and it compares the failure
mechanisms in isolated and strip anchors to determine theshape effect on the uplift resistance.

ABSTRACT: The progressive failure of a shallow anchor embeddedin granular materials was evaluated in terms of scale, embedment,
and shape effect by conventional 1 g tests and extensive elastoplastic
finite element (FE) analysis. The model tests were performed using a
trap door and isolated anchors in Toyoura and Soma sands. The FE
analysis employed a constitutive model for a non-associated strain
hardening/softening elastoplastic material. This model incorporated
the effect of shear band thickness. Progressive failure depended on
the density of the testing bed, h/D, and the anchor shape. The shapeeffect of an isolated anchor was considered to be due to progressive
failure.

Keywords: anchor, progressive failure, scale effect, shapeeffect, shear band
1. INTRODUCTION
Anchors are widely utilized for stabilizing towers, bridges,
slopes, and various other structures. Several studies haveinvestigated the uplift resistance of anchors [1–3]. Theoreticalstudies have focused on stability analysis based on rigid
plastic and elastoplastic FE analysis [4–6]. Rigid plastic
analysis is widely employed for designing structures due to its
convenience. However, it requires making arbitrary
assumptions regarding the shape of the failure surface
observed during model tests. If all tests are conducted in thesame granular materials with a constant density, the values of
Nq (= T/Aγh, where T is the uplift capacity, γ is the effective
unit weight of sand, h is the embedment depth, and A is theanchor area) obtained by rigid plastic analysis will remain
relatively constant for constant h/D (where D is the diameteror the smallest dimension of the anchor). De Beer [7]discussed the scale effect of the footing problem and showedthat the bearing capacity factor generally decreases withincreasing footing width. The scale effect is caused by the

2. TESTING APPARATUS
Tests were performed using an anchor pullout machine forisolated anchors and a trap-door testing apparatus for strip
anchors. The cylindrical container used in these tests was 59

Fig.1 Anchor testing apparatus

Fig.2 Trap-door testing apparatus
23

cm in diameter and 30 cm in height (Fig. 1). Flat circular andsquare anchors made of 0.5-cm-thick steel plate with widths(B) or diameters (D) of 5, 10, and 15 cm were used. The
rectangular anchors had aspect ratios (L/B, where L is thelength of the anchor) of 1, 2, 3, 4, 5 and 6. The trap-door
testing apparatus consisted of a glass-walled box and a trap
door. The testing apparatus was 70 cm long, 50 cm high, and40 cm wide (Fig. 2) and the trap door width (D) was 10 cm.
These dimensions were selected so that boundary effects
could be neglected. Granular masses were prepared bypluviating air-dried Toyoura and Soma sands, which hadmean particle diameters of 0.16 and 0.45 mm and relative
densities of Dr = 90, 60, and 45% and Dr = 91%, respectively.
The tests were performed for embedment ratios (h/D; h:granular mass depth) of 1, 2, and 3. All tests were conductedunder normal gravity conditions. To study the propagation of
the shear band, colored granular layers were placed against
the front glass wall of the apparatus.
3. NUMERICAL MODELS

band area to the FE area:

S

Fb
Fe

( 1)

where Fb is the shear band area in each FE and Fe is the FE
area. The shear band thickness was used based on the resultsof shear band thickness studies.
The yield function（ f ）and plastic potential function（  ）are given by the following expressions,f  3 ( ) m J2
  ( )  0g( )  3 ( ) m  J 2   ( )  0(3)p   d(4)
where,  m is the mean stress, J 2 is the second invariant ofdeviatoric stress and  is the Lode angle.
'

In 2D FE analysis, the frictional hardening/softening function
( ) is expressed asm2  f  ( )    p (hardening-regime;    f ) (5)   f     f 2 
( )  r  ( p  r )exp
   r  (softening-regime;    f )(6)
where, m ,  f and r are the hardening/softening material

In the FE analysis, the shear band effect was introduced in theconstitutive equation. An elastoplastic model with a
non-associated flow rule and strain hardening/softening
material properties was used in the constitutive model. A
Mohr–Coulomb yield function and a Drucker–Prager plastic
potential function were employed. Four-node Lagrangianelements were used with reduced integration. A dynamic
relaxation method with a return mapping algorithm wasapplied to the integration algorithm.
To include the mesh-size effect induced by softening of thematerial, a characteristic length scale was introduced. It is afunction of the shear band thickness and the ratio of the shear

parameters.
The parameters of  p and  r are estimated using thefollowing equations,
Table 1 Summarize the parameters used in the FE analysis

Fig.3 2D FE meshAn chor

Anc horFig.4 3D FE meshes



24







(2)

2sin  p
(7)
3(3  sin  p )2sin  r(8)
r 
3(3  sin r )where,  p and  r are the peak and residual friction angle,

In 3D FE analysis, simple hardening/softening functions wereused for 3D problems to save computational time. The
frictional hardening/softening function ( ) and  ( ) are
expressed as

p 

(17)
(18)

respectively.
The peak friction angle of  p is estimated from the empirical

(19)

√

relations proposed by Bolton (1986).

 
2
Ir  Dr 1 （  m  147kN/m ） (9)5  ln m 



150
2
(10)
Ir  5Dr 1 （  m  147kN/m ）(11)
 p  3I r   rwhere, Dr is relative density.
The plastic potential function () is defined as,
2sin
(12)
( ) 
3(3  sin )

1√

sin  mob  sin r(13)
1 sin  mob sin r  2 (14)
r   r 1  exp   d  where,  and d are the stress-dilatancy material

parameters.
The function    can be neglected in the case ofcohesionless soil.
The elastic moduli are estimated from modified the equation
proposed by Hardin(1968) and are given the following
equations in case of clean sand:

4. PROGRESSIVE FAILURE
Fig. 5 shows shear band development in a sand mass. These
tests were performed in dense Toyoura sand on a trap door
with D = 10 cm and h/D = 2. Two almost identical shear
bands developed from the edges of the trap door into the sandmass. As the trap door moved upward, outer primary shear
bands became inactive and additional steeper inner shear
bands developed. The shear bands propagated upward.
Fig. 6 shows the development of shear bands in Toyoura andSoma sands in a semipartial testing apparatus at adisplacement of 5.0 mm. Since similar shear bands developed
from the edges of the trap door, photographs were taken fromthe left-hand side. A secondary steeper shear band developed



m

(15)

1 e
1(1  )(16)KG3(1 2 )where,  is Poisson’s ratio， e is the void ratio and G 0 is

the constant of initial shear modulus.



(22)

The values of  f , r ,  d , m , B and C were evaluated on thesin 

0(21)

back-prediction of a triaxial compression test by comparing
results of finite element analysis with experiment. Table 1
summarizes the parameters used in the finite element analysis.
The details of the numerical method were shown in referenceof Sakai and Tanaka [8,9].
Fig. 3 shows the FE mesh used for the 2D analysis. The
failure mechanism in the rectangular anchor foundation iscompared with those in the square and strip anchors todetermine how the shape effect affects the uplift resistance. In
this regard, 3D models with H/D = 2, D = 50 mm, and L/D = 1,
2, 3, 4, 5, 6, 7, 8, 10, and 20, and a 3D plane strain model (L/D= ∞) were created to investigate how the shape effect affects
the uplift resistance of the rectangular anchors. Some meshes
are shown in Fig. 4.

The dilatancy angle of  is estimated from modified Rowe’s
stress-dilatancy relationship,2
2.17  e  0.5

GG

(20)

√



(a)Disp.=4mm(b) Disp.=8mm(c)Disp.=12mm
Fig.5 Shear band development in trap-door test (D=10cm, h/D=2, dense)25

in Toyoura sand after the outer primary shear band became
inactive. Shear band propagation was greater for smaller
grain diameters. Toyoura and Soma sands had shear band
thicknesses of about 3 and 9 mm, respectively. The observed
shear band thicknesses were about 20 times greater than themean grain diameter. Fig. 7 compares the uplift resistanceagainst displacement curves obtained using Toyoura sandwith different densities and a circular anchor with D = 5 cm.
For high-density sand, the uplift resistance clearly decreases
after the maximum uplift resistance is attained and obvious
softening occurs. For loose conditions, the uplift resistanceremains almost constant after the maximum uplift resistanceis realized and softening is not obvious. The residual uplift
resistances were similar for all densities. Fig. 8 shows shear
band propagation at a displacement of 10 mm for dense andmedium conditions. Shear band propagation progressed morein dense conditions. Fig. 9 compares the Nq and displacement
curves (D = 5 and 10 cm; h/D = 1 and 3) obtained using acircular anchor. For the embedment ratio, h/D = 3, a
remarkably large displacement difference is required to attainpeak Nq between D = 5 and 10 cm. Remarkable progressive
failure occurs for h/D = 3.

(a) Dense(b) LooseFig.8 Shear band development in dense and loose
condition (Toyoura sand, D=5cm, Disp=10mm)

Fig. 9 Nq-displacement curves obtained by experiments(Dense)5. SCALE EFFECT(a)Toyoura Sand(b) Soma Sand
Fig.6 Shear band development in sand mass
(D=5cm, Disp.=5mm)

Fig. 10 compares the numerically and experimentallyobtained uplift resistance–displacement curves in loose,medium, and dense beds for a circular anchor. These curves
were obtained for a circular anchor with Toyoura sand for D =10 cm and h/D = 2. These results reveal that the numericallycalculated maximum uplift resistance, displacement
corresponding to the maximum uplift resistance, and residual
uplift resistance coincide with those obtained fromexperimental results. Figs. 11 and 12 show that the shear bandpropagates upward from the edge of the anchor plate and thatthe directions of the narrow localized zones obtained from the
distributions of the maximum shear strain in FE analysis
almost coincide with the directions of the outermost shear
bands observed experimentally for loose, medium and dense
sands. Fig. 13 shows the relationship between the peak Nq
and D obtained by experiment and numerical analysis for
dense, medium, and loose beds. Both experimental andnumerical results show that the peak Nq generally decreases
with increasing D in the dense bed and that peak Nq does notdecrease in the loose bed. The scale effect was clearlyobserved in the dense bed. Fig. 14 shows the relationshipFig.7 Uplift resistance-displacement curves on circularanchor obtained by experiments (Toyoura sand, D=5cm)

26

Fig.10 Uplift resistance-displacement curves on circular anchor obtained by experiment and numerical (D=10cm)(a) loose(b)medium(c)dense
Fig.11 Experimental shear band development on circular anchor (Toyoura sand, D=5cm, Disp.=from 4mm to 5mm)(a) loose(b) medium(c) dense
Fig.12 Maximum shear strain distribution on circular anchor (D=5cm, Disp=2.5mm)

experimental and numerical results agree closely with eachother. Fig. 17 shows the relationship between the mobilized
peak resistance factor (Nq) and L/B. It decreases rapidly withincreasing L/B when L/B < 6 and it decreases gradually as
L/B increases above 6 to the strip anchor condition (L/B = ∞).
Fig. 18 shows that the outermost shear band in the numericalanalysis outcropped on the ground surface for L/B = 1 and
L/B=6. Fig. 19 shows a characteristic photograph of theground surface for L/B = 1. The failure mechanism of theshallow anchor is associated with the progressive formationof a shear band. To clarify the failure mechanism of therectangular anchors, Fig. 20 shows a plot of the numericalmaximum shear strain distribution in rectangular (L/B = 6),
strip (L/B = ∞), and square (L/B = 1) anchors at
displacements of δ/D = 0.016 (at the peak) and 0.04 (after the
peak uplift load). The shear strain distribution is not uniform
in the propagating upward shear bands in the anchor. It is
slower or less progressive at the corners of the square anchor

between the peak Nq and D obtained by experiment andnumerical analysis for h/D = 1, 2, and 3. Both theexperimental and numerical results show that the peak Nqgenerally decreases with increasing D. The reduction in thepeak Nq with increasing D becomes remarkable withincreasing h/D. The scale effect was remarkable for h/D = 3.
Fig. 15 shows the relationship between the peak Nq and D
obtained experimentally and by numerical analysis for strip
and circular anchors. The results for both anchors exhibit ascale effect, but it is particularly remarkable for the circularanchor.
6. SHAPE EFFECT
The curves in Fig. 16 depict the experimental and numerical
relationships between the pullout resistance and thedisplacement factor for the square and rectangular anchorswith L/D = 1, 2, 3, 4, 5, and 6 (D = 50 mm and h/D = 2). The
27

DenseMediumLoose

Fig.13 Peak Nq –D-density relationshipFig.16 Relative displacement-uplift resistance curves ofrectangular anchor in dense sands

Fig.14 Peak Nq –D-h/D relationship in dense sands

Fig. 17 Peak Nq –L/B relationship in dense sands

Stra ight
Ci rcular arc(a) L/B=1 (square)(b)L/B=6
Fig.18 Outermost shear strain distribution in dense sands

Fig.15 Peak Nq -D relationship(strip and circular anchor, Dense)
Fig.19 Deformed groundobtained by experimentthan in the strip anchor region of the rectangular anchor. Thus,
as L/B increases, the strip-type mechanism starts to dominate
over the square-type mechanism. Consequently, the softening
28

rate decreases after the peak uplift resistance factor is
obtained and the mobilized peak resistance factor decreases
with increasing L/B.

Strip (L/B= ∞)Rectangular (L/B=6)

Square (L/B=1)

cut&
op enedcut&
open edcut&o penedδ/B =0.016

δ/B=0.016

δ/B=0.016

δ/B=0.04

δ/B =0.04

δ/B=0.04%

Fig.20 Maximum shear strain distribution obtained by 3D FE analysis (Dense)[5]

Walters, J. V. and Thomas, J. N. “Shear zone development in granularmaterials,” Proc. 4th Int. Conf. on Numerical Method in Geomech,pp.263-274, 1982
[6] Rowe, R. K. and Davies, E. H. “The behavior of anchor plates in sand,”
Geotechnique, 32, pp.25-41, 1982
[7] de Beer, E. E. “Bearing capacity and settlement of shallow foundations
on sand,” Proc. of a Symp. held at Duke University, 1965
[8] Sakai, T. and Tanaka, T. “Scale effect of a shallow circular anchor indense sand,” Soils and Foundations, 38(2), pp93-99, 1998
[9] Sakai, T. and Tanaka, T. “Experimental and numerical study of uplift
behavior of shallow circular anchor in two-layered sand,” J. of
Geotechnical and Geoenvironmental Engineering, 133(4), pp.469-477,
2007[10] Vermeer, P. A. and Sutjiadi, W. “The uplift resistance of shallow
embedded anchors,” Proc. 11th Int. conf. on Soil Mechanics andFoundation Engineering, 4, pp.1635-1638, 1985
[11] Murray, E. J. and Geddes, J. D. “Uplift of anchor plates in sand,” J. of
Geotechnical Engineering, 113, 202-215, 1987
[12] Merifield, R. S. and Sloan, R. S. “The uplift pullout capacity of anchorsin frictional soils,” Can. Geotech. J. 43, pp.852-868, 2006[13] Ovesen, N. K. “Centrifuge tests of the uplift capacity of anchors,” Proc.
7th European Conf. Soil Mech. And Found. Eng., 4, pp.318-323, 1981[14] Dickin, E. D. “uplift behavior of horizontal anchor plates in sand,” J. of
Geotechnical Engineering, 114, pp,1300-1317, 1988
[15] Vardulakis, I., Graf, B. and Gudehus, G. “Trap-door problem with dry
sand; a statistical approach based upon model test kinematics,” Int. J.
Numer. Anal. Meth. Geomech., 5, pp.57-78, 1981[16] Yoshida, T., Tatsuoka, F., Siddiquww, M. S. A., Kamegai, Y. and Pak,
C. “Some observation of zone of localization in model tests on dry
sand,” Proc. 3rd Int. Workshop on Localisation and Bifurcation Theory
for Soils and Rocks, pp.165-180, 1993[17] Sakai, T. and Tanaka, T. “Uplift behavior of circular and square anchor
foundations in dense sand,” Doboku Gakkai Ronbunsuu C, 65(1),pp131-137, 2009

7. CONCLUSION
This study sought to explain the progressive failure ofshallow anchors in granular materials by performing
conventional 1 g tests and extensive numerical studies. The
conclusions can be summarized as follows:
1) Shear band propagation was investigated for various
anchor problems. Progressive failure was found to depend on
the testing bed density, h/D, and the anchor shape. The scaleeffect of the anchor in dense sands is considered to be due toprogressive failure. It was not clearly observed in loose beds,
for small h/D, or for strip anchors.
2) The failure mechanisms of rectangular and strip anchors
were compared to determine the effect of the shape effect on
the uplift resistance. The shape effect is considered to be due
to progressive failure. The softening rate decreases after the
peak uplift resistance factor is obtained and the mobilized
peak resistance factor decreases with increasing L/B. It is thus
necessary to use 3D models to model the effects ofgeometrical factors.

8. REFERENCES
[1]
[2]
[3]
[4]

Balla, W. “The resistance to breaking out of mushroom foundations forpylons,” Proc. 5th Int. Conf. on Soil Mechanics and Foundation
Engineering, pp.569-576, 1961
Meyerhof, G. G. and Adams, J. I. “The ultimate uplift capacity offoundations,” Can. Geotech. J. 5, pp.225-244, 1968
Matsuo, M. “Study on the uplift resistance of footing (1),” Soils anf
Foundations, 7, pp.1-37, 1968
Davie, J. R. and Sutherland, H. B. “Uplift resistance of cohesive soils,”
J. Geotec. Engrg. Div. 103(9), pp.935-952, 1977

29

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Design of Single Pile Foundations Using the Mechanics of Unsaturated Soils
Vanapalli S.K., Civil Engineering, University of Ottawa, Ottawa, Canada
Taylan Z.N, Civil Engineering, Istanbul Technical University, Istanbul, Turkeyconventional techniques used in the design of piles ingeotechnical engineering practice. These methods arepresented in a functional form such that they can be used for
estimating the variation of the shaft capacity of the single piles
with respect to matric suction using the saturated soil
properties and the Soil-Water Characteristic Curve (SWCC).
The proposed modified equations take the conventional formof the  and  methods used for saturated soils when thematric suction value is set to zero.
A series of single model pile tests placed in staticallycompacted unsaturated glacial till with various degrees ofsaturation were performed in a laboratory environment tostudy the influence of matric suction on the shaft resistance.
The results of these experimental studies were interpretedusing the modified , and methods. There was areasonably good comparison between the measured ultimate
shaft capacity of the single model piles and those estimatedusing the proposed methods (i.e., modified  and 
methods).

ABSTRACT: The load carrying capacity of single piles arecommonly estimated using the well known  [1],  [2-3] and [4]methods. These methods are also used in engineering practice forunsaturated soils even though the ,  and  methods are based onconventional saturated soil mechanics. In this study, a series of single
model pile tests were conducted in a laboratory environment to study
the influence of matric suction on the pile shaft capacity in astatically compacted fine-grained soil. The results of the study showthat the shaft capacity of single piles is significantly influenced by thecontribution of matric suction. Based on the experimental results, theconventional ,  and  methods were modified to estimate the total
shaft resistance of piles in unsaturated soils by including theinfluence of matric suction. The modified methods can also be used
for estimating the variation of shaft capacity of single piles withrespect to matric suction using the Soil-Water Characteristic Curve(SWCC) and the saturated shear strength parameters. The modified , and  methods are promising for use in engineering practice toestimate the ultimate shaft bearing capacity of single piles placed inunsaturated soils.

Keywords: Unsaturated soils, pile design, modified , and 
methods, SWCC, matric suction

2 BACKGROUND
1 INTRODUCTION
The reliable determination of soil-structure interaction
parameters requires cumbersome laboratory or field tests.
Alleviating the need of such cumbersome tests; empiricalmethods are proposed to estimate the skin friction, fs based on
the conventional shear strength parameters and theinformation related to the variation of effective stresses along
the length of the piles.

In geotechnical engineering practice, conventional soil
mechanics principles are used for the design of pilefoundations assuming the soil is in a state of saturated
condition [5], [6]. However, in many situations, the groundwater table is at a great depth and natural soils are typicallyfound in a state of unsaturated condition. This is particularly
true for soils in arid and semi-arid regions. Several
geotechnical structures such as highways, embankments,
dams are constructed on or with compacted unsaturated soils
in which pile foundations may be placed. The stresses
associated with these foundations are distributed within theunsaturated soil zone above the ground water table.
Several recent studies on shallow foundations [7]-[10] haveshown that the bearing capacities of both coarse andfine-grained soils are significantly influenced due to thecontribution of matric suction. However, limited numbers ofstudies are reported in the literature [11]-[14] that consider the
influence of matric suction or capillary stresses on the load
carrying capacity of pile foundations. Typically, pilefoundations are designed assuming saturated, dry orsubmerged soil conditions.
In this paper, the  method by Skempton [1],  method byChandler and Burland [2], [3] and  method by Vijayvergiya
and Fotch [4] are modified such that they can be used for
estimating the ultimate shaft resistance of piles in unsaturated
soils. The modified , and  methods are similar to thef s  f  v ,  , cu 

(1)where, v = vertical effective stress, '= effective friction
angle, and cu = undrained shear strength.
The functional form of (1) suggests that the skin friction, fs
can be analyzed in terms of either total or effective stressapproach considering the loading and drainage conditions(i.e., TSA or ESA), respectively.
Experimental programs were planned to determine thecontribution of matric suction on the shaft resistance and not
the bearing resistance. In other words, the contribution of end
bearing capacity is not measured in the present study.
2.1 The  Method (ESA)In both coarse and fine-grained soils, the skin friction, fs
mobilized along the length of the pile is a key parameter that isrequired in the estimation of the load bearing capacity of pilefoundations. If a pile is loaded at a relatively slow rate (i.e., to31

The adhesion factor,  can be computed from American
Petroleum Institute charts which are given as a relationshipbetween the adhesion factor and the undrained shear strength.
Alternatively, (6) can also be used for estimating the  value.

achieve drained conditions) the skin friction resistance can be
estimated using (2) [2].

f s  c  K0 v tan  '(2)where, K0 = mean lateral earth coefficient at rest, c' and ' =effective cohesion and internal friction angle of soil,
respectively, and ’v = vertical effective stress along the pilelength.
The shear strength of soils associated with cohesion decreases
significantly due to the remolding and softening effects duringpile installation. This leads to an assumption that effective
cohesion can be neglected along the pile shaft, particularly incoarse-grained soils and other soils with low percentage offines such as silty sands and normally consolidated clays.
Hence, (2) can be incorporated in (3) with the introduction of acoefficient, β [2], [3].
Q f  f s As   v  d L

  0.5 0.5 if   1
  0.5 0.25 if   1
where  = cu/v and v = vertical effective stress [17].
2.3 The  Method

The conventional  method combines the total (i.e.,undrained) and effective (i.e., drained) stress approaches forcalculating the shaft capacity of piles driven into fine-grainedsoils [4]. This technique is useful in reducing the sensitivity ofthe shear strength parameters measured using the TSA and
ESA. The total shaft capacity is calculated using therelationship shown in (7).

(3)

where, β is a coefficient which is equal to Ko tan δ’, δ’=effective angle of friction along the soil/pile interface, As =surface area of the pile, ’v = vertical effective stress at the
mid of the pile shaft, L = length of pile, and d = diameter ofpile.
The coefficient, β values typically vary from 0.30 to 0.60 forfine and coarse-grained soils [2], [3].



(7)where, σ'v(avg) = the mean effective stress, cu = undrained shear
strength along the pile length,  = frictional capacity
coefficient which is a function of entire embedded depth of
pile. The coefficient  varies from 0.12 to 0.5 for pilepenetration of 0 to 70 m based on the 42 piles load test data
gathered and presented by [4].

Undrained lading conditions can be assumed when a pile isloaded at a relatively fast rate in saturated fine-grained soils.
The ultimate shaft resistance can be estimated for such loading
conditions extending the TSA. In other words, the ultimate
shaft capacity of a pile, Qf is dependent on the undrained shear
strength, cu of the soil. Hence, the unit skin resistance, fs canbe expressed as (4) using undrained shear strength, cu and theadhesion factor, .

3

ESTIMATION OF THE ULTIMATE SHAFTCAPACITY (USC) OF PILES IN UNSATURATED
SOILS

3.1 Modified β method

Researchers ([18], [19]) proposed a model (8) to predict thevariation of shear strength with respect to matric suction using
the SWCC and the effective shear strength parameters (i.e., c'and ') as given below.

(4)

where,  = adhesion factor between soil and pile.
There is a large data base of in-situ pile load tests includingbored and driven piles supporting this method dating back to
1950s. This method is commonly referred in the literature as
the  method. The studies show that adhesion factor,  is notconstant but decreases with increasing undrained shear
strength, cu of the soil and varies from close to unity for low
strength soft clays and reach almost to a value of 0.4 for stiffclays for cu values greater than150 kPa [16], [1]. The ultimate
shaft capacity, Qf for cylindrical piles using the  method canbe estimated as (5).
Q f  f s  As   cu  d L

Q f    v' ( avg )  2cu  dL

2.2 he  Method - (TSA)f s   cu(6)

  c   ( n  ua ) tan     ua  uw  ( S  )(tan  ) (8)where, c' = effective cohesion, (σn - ua) = net normal stress,
' = effective internal friction angle, (ua - uw) = matric suction
S = degree of saturation,  = fitting parameter used for shear
strength.
The contribution of matric suction towards the shear strength,us can be expressed as (9) which is the second part of (8).
(9)
 us   ua  uw  ( S  )(tan  ) (5)The contribution of matric suction, towards the ultimate shaft
capacity of a single pile, Q(ua-uw) can be estimated using (10) as
given below.

where, d = pile diameter, and L = length of pile.

32

Q(ua  uw )   us As   ua  uw  ( S  )(tan  )   d L

where, cu(sat), and cu(unsat) = undrained shear strength undersaturated and unsaturated conditions, respectively, Pa =atmospheric pressure (i.e. 101.3 kPa), and , and  = fitting
parameters.
The fitting parameter  is dependent on the soil type (i.e.,coarse or fine-grained soils) and is equal to 1 forcoarse-grained soils and 2 for fine-grained soils. The fitting
parameter  however is a function of plasticity index, Ip.

(10)Equation (10) suggests that the variation of ultimate shaft
capacity with respect to matric suction can be estimated using
the SWCC and effective interface friction angle, δ'.
A general expression for estimating the ultimate shaft capacityof piles in unsaturated soils, Qf(us) can be obtained by
combining (3) and (10) as given below 9Q f (us )  Q f ( sat )  Q f (ua uw )   v   ua  uw  ( S  )(tan  )   d L  2.1088 e(11a)

Q f (us )  ca    z    ua  uw  (S )(tan  )   d L(ua  uw ) Q f (us )   cu ( sat ) 1 
S /    dL
 ( Pa / 101.3)
The  method was modified to propose (15) to include theinfluence of matric suction in the estimation of shaft resistanceof piles in unsaturated soils.

S  /  P/101.3
 a



 '





 (Pa /101.3)

Qf (us)   v(avg)  2cu(sat) 1

(11b)(ua  uw )S /   d L




(15)

The form of (15) will be as same as the conventional  method
once the matric suction is set to zero. Equation (15) can also
be used to estimate the variation of total shaft resistance of
pile, Qf(us) with respect to matric suction. The required
information for (15) are the undrained shear strength undersaturated condition, cu(sat) and the SWCC. More details of thismethod are discussed while analyzing the results.

Several investigators related the load bearing capacity of asingle pile to the undrained shear strength, cu of thefine-grained soils ([16], [21]-[28]). In the present study, theconventional method is modified such that it can beextended for interpreting the results of model piles tested
under unsaturated soil condition. In addition, a model is
proposed for estimating the variation of ultimate shaft
capacity of model piles with respect to matric suction.
The equation (12) [9] can be used to estimate the variation ofundrained shear strength with respect to matric suction using
the SWCC and undrained shear strength for saturated
condition, cu(sat).

 ua  u w (14)

3.3 Modified  method

3.2 Modified  method



(13)

The undrained shear strength under saturated condition, cu(sat)and the SWCC are required to estimate the variation ofultimate shaft capacity of pile, Qf(us) with respect to matricsuction. Equation (14) will be the same as (4) when the matricsuction value is set equal to zero.

where, ca' = adhesion component of cohesion for saturated
condition, δ'= effective angle of interface along the soil/pile.

cu (unsat )  cu ( sat ) 1 

(15.5  I p (%)  60)

Following the procedure described in section 3.1, the ultimate
shaft capacity of piles in unsaturated soils under undrained
loading conditions can be estimated by combining (4) and (12)as given below.

The relationship between the fitting parameter,  andplasticity index, Ip provided for predicting the shear strengthof unsaturated soils [20] can be used for estimating theultimate shaft capacity of a single pile. More details of thismethod are available in [14]. Equation (11a) shows that there
is a smooth transition between the ultimate shaft capacity of asingle pile from an unsaturated to saturated condition. This
relationship will be the same as (3) when matric suction isequal to zero (i.e., for saturated soils).
The contribution of cohesion component associated with theadhesion, ca' under drained loading condition may not be
negligible for evaluating the pile capacity of fine-grained soils
for the method (see (11b). In other words, there will be some
contribution of adhesion, ca' towards the ultimate shaft
capacity which will be mobilized with time after the
installation of the pile. Therefore, the ultimate shaft capacityof piles in unsaturated fine-grained soils under drained
loading conditions can be estimated by modifying (11a) as
given below.
(8.0  I p (%)  15.5)
0.0903 ( I p )4 TESTING PROGRAM
A series of model pile load tests were performed in saturated
and unsaturated compacted fine-grained soil under drained
and undrained loading conditions. The soil chosen for thisstudy is a glacial till obtained from Indian Head,
Saskatchewan, Canada. The key objective of the test programis to determine the influence of matric suction on the ultimate
shaft capacity of model piles.

(12)



33

4.1 Soil Properties
The properties of the tested soil are summarized in Table I.
Procedures followed for determining some of the test resultssummarized in Table I are not detailed in this paper due tospace limitations. The shaft bearing capacity of model piles
were proposed to be determined at three different watercontents; 13% (dry of optimum), 16% (dry of optimum) and
18% (close of optimum). These water contents were chosenfrom the compaction curve data. The dry densities of thecompacted soil at these water contents were respectively equal
to 14.5 kN/m3, 16.1 kN/m3 and 16.7 kN/m3. The matricsuction values of the tested compacted soils were measured
using the axis-translation technique [29] with a modified nullpressure plate [30]. The measured matric suction values were
205 kPa, 110 kPa and 55 kPa for water contents of 13%, 16%,
18% respectively. The experiments were not conducted for thewater contents on the wet of optimum side since the degree ofsaturation values were greater than 90%, which resulted insignificantly low matric suction values.
The measured SWCCs of the specimens prepared with initialwater contents of 13%, 16% and 18% are presented in Fig. 1.
These SWCCs were measured following the drying path using
the pressure plate apparatus. More details of the matric suction
measurements using both the modified null pressure plate andthe SWCC using pressure plate are discussed in [31], [32].

Degree of Saturation, S (%)

100

Optimum water content, wopt (%)
Maximum dry unit weight, γdmax (kN/m3)
Saturated unit weight, γsat (kN/m3)
Sand (%)
Silt (%)
Clay (%)
Liquid limit, LL (%)
Plastic limit, PL (%)
Plasticity index, Ip (%)
Interface friction angle (soil/pile)' (deg.) (Sat/Unsat)
Apparent cohesion, c'a (kPa) (Sat/Unsat)
Effective cohesion, c' (kPa) (Sat)
Effective friction angle, ' (deg.) (Sat)
Undrained shear strength, cu (kPa)

80
70
60
50

winitial =13%
winitial =16%40

winitial =18%

30
1

10

100

1000

Matric Suction, (ua- uw) (kPa)

Figure 1. SWCCs for the Indian Head till prepared at threedifferent initial water contents.
The soil-water mixture (hereafter referred to as soil) was
placed in a tank (300 mm in diameter and 300 mm in height).
The soil was compacted statically with 350 kPa stress into thetest tank using a specially designed compaction base plate.
The compaction and model pile load tests were conductedusing a conventional triaxial test loading frame (Fig. 2).

Table I. Properties of the tested soil
Soil Properties

90

Indian Head
till
18.6
16.7
18.5
28
42
30
32.5
17
15.5

①
④
③

②

⑥

⑤

25/27
20/100
15
23
11.5

②

Figure 2. Test setup for model pile loading test: ① Adjustableheight loading frame ② Test tank ③ LVDT ④ Load cell ⑤ Modelpile, ⑥ Compaction base plate.

4.2 Testing Methodology and Equipment Details
Special testing procedures were followed to determine theultimate shaft capacity of the model pile placed in thestatically compacted soil. The model piles were loaded insaturated/unsaturated compacted soils for both drained andundrained loading conditions. The soil sample collected fromthe field was air-dried for several days, subjected to gentle
pulverization, passed through a sieve with an opening size of 2
mm (i.e., #10 sieve), and oven-dried.
The oven dried soil, after reaching the room temperature in thelaboratory, was mixed with distilled water at predetermined
initial water contents. The prepared soil-water mixture wasplaced in sealed double plastic bags and then stored in ahumidity controlled box for at least 3 days to ensure uniform
water content conditions throughout the sample.

After the soil was compacted under static loading conditions
in five layers in the test tank, a thin wall sampling tube of 18.7
mm diameter with 1 mm of wall thickness was used to create ahole down to a depth of 220 mm. The sampling tube alongwith the soil column embedded into it was removed out of thecompacted soil. The model pile used in the study was madeout of stainless solid steel cylindrical rod with 20 mm
diameter. The model pile (hereafter referred to as D-20 pile)was slightly larger in diameter in comparison to the diameterof sampling tube in order to obtain a good contact between thewalls of the drilled shaft and the model pile. After the borehole
drilling was completed, model pile was jacked down to adepth of 200 mm. A gap with 20 mm of length at the tip of the34pile was intentionally left to eliminate the end bearing
resistance while loading the model pile. In other words, thevoid was created to facilitate in the measurement of shaft
resistance without any contribution from the end bearing
resistance. The tests described were conducted underunsaturated (UNSAT) conditions.
However, when the pile was loaded under saturated (SAT)
condition, the compacted unsaturated soil was graduallysaturated by allowing downward flow of water from the top of
the soil through the compaction base plate which had
apertures. The compactor plate was placed on top of thecompacted soil sample and fixed to the loading frame in order
to avoid possible volume change due to swelling. After thesaturation process was completed, the model pile was installedusing same procedure described for the pile testing underunsaturated condition.
A piezometer attached to the side of the tank was used tocheck the saturation condition. The soil was assumed to besaturated as the level of water in the piezometer reached thesame water level within the test tank. The degree of saturation
was also verified by measuring matric suction with atensiometer that was placed in the compacted soil. The
tensiometer reading of (ua – uw) = 0 kPa provided anotherindication that the compacted soil is saturated. In addition tothese checks, small chunks of soil specimens were collectedfrom the tank for water content measurements after theloading tests were completed. The average water content fromthese tests was 31% which corresponds to a degree ofsaturation equal to 96% calculated from mass-volume
relationships. This value can be considered to be close tosaturation conditions for the present study because there isother evidence to support the soil is saturated.
A strain rate of 0.0120 mm/min was chosen for loading thepile to achieve drained loading conditions [18][33]. A
relatively fast loading rate of 1 mm/min was used to simulate
undrained loading conditions [34].
The soil samples prepared with an initial water content of 13%
were tested under both unsaturated and saturated conditions
considering both the drained and undrained loadings (i.e.,
SAT-Drained, SAT-Undrained, UNSAT-Drained, UNSATUndrained). Only the tests under unsaturated conditions wereconducted (i.e., UNSAT-Drained, UNSAT-Undrained) for
other soil samples with initial water contents of 16% and 18%.
The details of the experimental program of the present study
are summarized in Fig. 3 as a flow chart.

are similar to the results published on results on otherfine-grained clays in the literature [1],[3].
Figure 3. Flow chart of the testing program.
Sample preparation(Indian Head till, 2mm sieve,
winitial = 13-16-18%

Load test
UNSAT Drained
winitial =13-16-18%(ua - uw) = 205,110,53 kPa

Static compaction(350kPa, five layers)
Load test
UNSAT Undrained
winitial =13-16-18%(ua - uw) = 205,110,53 kPa

Saturation(Downward flow of water)

Load test
SAT Drained
winitial = 13%(ua - uw) = 0kPa

Load test
SAT Undrained
winitial = 13%(ua - uw) = 0kPa

5.2 Interpretation of the Test Results
The measured shaft bearing capacity values for the D-20 pile
were interpreted using the modified ,  and methods
proposed in this paper. In addition, comparisons are provided
between the measured and estimated shaft bearing capacity
values.
The ultimate shaft bearing capacities for the model piles wasestimated using the modified  method assuming undrained
loading conditions. The undrained shear strength values forthe compacted soils required for the modified  method were
determined by conducting unconfined compression tests on
the samples collected from the testing tank after model pile
tests. The undrained shear strength values were also estimatedusing the equation (12). The test results are summarized inTable II.
Table II. Comparison between the measured and estimated ultimate
shaft capacities using the modified method for undrained loading.

5 MODEL PILE TEST RESULTS AND ANALYSES
5.1 Test Results
The model pile test results obtained for both saturated andunsaturated soil samples are presented in Fig. 4 through Fig. 7.
The shaft carrying capacity of the model piles loaded in thesoils compacted with different compaction water contents (i.e.,
13%, 16%, and 18%) is significantly different depending on
the soil (saturated or unsaturated) and drainage (drained orundrained) conditions. The trends of the load versus
displacement behavior of model piles from the present study
35

Back
Cal.
 value

Est.4
Qf(us)Meas.
Qf(us)

(%)
kPa
kPa
kN
13
0
11.5
0.90
0.70
0.13
18
53
58
52
0.82
0.68
0.59
13
205
68
57
0.75
0.79
0.64
16
110
80
65
0.67
0.55
0.67
(ua - uw) = matric suction
1
Undrained shear strength from unconfined compression tests.
2
Undrained shear strength calculated by using (12)3value obtained using the correlation charts [24]4Calculated shaft bearing capacity by using (14).

kN
0.10
0.50
0.68
0.55

winitial(ua - uw)

cu1

cu 2

3

0.6

Shaft bearing capacity (kN)

Shaft bearing capacity (kN)

1.0

0.8

0.6

0.4
UNSAT-Undrained
winitial = 13%

0.2

UNSAT-Drained
winitial = 13%

0.0
0.0

0.5

1.0

1.5

0.5
0.4
0.3
0.2
UNSAT-Undrained
winitial = 18%

0.1
0.0
0.0

2.0

0.5

1.0

Shaft bearing capacity (kN)

2.5

The measured ultimate shaft capacity and those estimatedusing the conventional (i.e. (2)) modified (i.e., (11b)) method for the saturated and unsaturated soils, respectively
under undrained conditions are summarized in Table III. The
fitting parameter  in (10) was determined using the relation
given in [20]. The coefficient,  = 0.3 was used for both thesaturated and unsaturated soils based on the soil-pile interface
friction angle,  (see Table III) since the influence of matricsuction on ’ is relatively less.

0.20

0.15

0.10
SAT-Undrained
winitial = 13%

0.05

Table III. Comparison between the measured and estimated ultimate
shaft capacities using the modified  method for drained loading.

SAT-Drained
winitial = 13%

0.00
0.0

0.5

1.0

1.5

winitial(ua - uw)





c'a

Est.
Qf(us)Meas.
Qf (us)

%
13

kPa
0

0.3

°
25

kPa
20

kN
0.25

kN
0.16

13

205

0.3

27

100

0.60

0.80

2.0

Displacement (mm)Figure 5. Model pile test results in a saturated soil compacted at an
initial water content of 13% under drained and undrained loading
conditions.

The measured ultimate shaft capacity values and thoseestimated using the modified  method (i.e. (15)). Vanapalli
and Taylan [35] analyzed the data available in the literature[4] and suggested the relationship between  and the ratio of
pile diameter to pile penetration depth, d/L are summarized inTable IV. A value of  = 0.32 was used in the present study.
More details with respect to using this value are detailed in
[35].

0.6

Shaft bearing capacity (kN)

2.0

Figure 7. Model pile test results in an unsaturated soil compacted at
an initial water content of 18% under undrained loading conditions.

Figure 4. Model pile test results in an unsaturated soil compacted at
an initial water content of 13% under drained and undrained loading
conditions.

0.5
0.4
0.3
0.2

Table IV. Comparison between the measured and estimated ultimate
shaft capacities by using the modified  method
UNSAT-Undrained
winitial = 16%

0.1
0.0
0.0

1.5

Displacement (mm)Displacement (mm)

0.5

1.0

1.5

2.0

2.5

3.0

Displacement (mm)Figure 6. Model pile test results in an unsaturated soil compacted at
an initial water content of 16% under undrained loading conditions.

winitial(ua - uw)

Meas.
cu



Est.
Qf(s),(us)Meas.
Qf(s),(us)

(%)
13
18
13
16

kPa
0
53
205
110

kPa
11.5
58
68
80

0.32
0.32
0.32
0.32

kN
0.09
0.47
0.55
0.64

kN
0.10
0.50
0.68
0.55

The measured and estimated results are summarized andpresented in Fig. 8. Each circle shown in Fig. 8 represents the36results of the three different methods with different matricsuction and initial water contents. The difference between themeasured and estimated shaft bearing capacities in terms ofpercentage varies between 6 to 36%. The difference is moresignificant for the results obtained using the modified βmethod (11b). Such a behavior can be attributed to the effectof loading rate and also due to the difficulties associated withthe opening a hole using the thin wall tube during which some
disturbance may have occurred.

experimental problems. Instead of drilling a hole into theinitially compacted soil, it is suggested to compact the soil
around the pile using a specially designed compactor. Such atechnique would reduce disturbance during pile installation
for tests and likely provide better correlation results as it
eliminates the problems associated with the soil disturbance.
In addition, this technique also provides better contact
between the pile and the soil.
The results summarized in this paper are based on the studies
undertaken using one compacted fine-grained soil. Moreexperimental and numerical studies are in progress to checkthe validity of the modified  and  methods for different
coarse and fine-grained soils. These studies will be useful tobetter understand the influence of matric suction on the load
carrying capacity of the piles. The results of the studies
conducted to date are promising for using the modified , and  methods in engineering practice to estimate the ultimate
shaft bearing capacity of single piles placed in unsaturated
fine-grained soils.

Measured shaft bearing capacity (kN)

1.2

1.0

0.8

measured = estimated
Modified method
Modified  method
Modified  method

winitial = 13%(ua - uw) = 110 kPa

0.6

winitial = 18%(ua - uw) = 53 kPa

7 ACKNOWLEDGMENTS

0.4

0.2

0.0
0.0

The first author thanks NSERC, Canada for providing funding
support in the form of Discovery Research Grant. The second
author thanks The Scientific and Technological ResearchCouncil of Turkey (TUBITAK) for providing the financial
support during this research. The authors would like to thankJean Claude Celestin (Technical Officer, Geotechnical Lab,
University of Ottawa), Won Taek Oh, Ph.D. candidate,
University of Ottawa and Mudasser Mohamed Anis Noor andNicholas Tang, undergraduate students of University ofOttawa for their time, help and assistance in the testing
program. Thanks also go to Hana Adem and Won Taek Oh,
Ph.D candidates, University of Ottawa for thier assistance inproof reading and editing this paper.

winitial = 13%(ua - uw) = 0 kPa

0.2

0.4

0.6

0.8

1.0

1.2

Estimated shaft bearing capacity (kN)

Figure 8. Measured and estimated ultimate shaft bearing capacities
calculated by using the modified ,  and  methods for compacted
Indian Head till samples

6 SUMMARY AND CONCLUSIONSThe conventional  and  methods are commonly used inengineering practice for estimating the ultimate shaft bearing
capacity of single piles in saturated soils. In the present study,
these methods are modified such that they can be used toestimate the variation of ultimate shaft capacity of single piles
with respect to matric suction using the SWCC and theconventional shear strength parameters. There is a smoothtransition between the modified and conventional methods
and are convenient to estimate the shaft capacity of piles inunsaturated and saturated soil conditions.
A series of model pile load tests were conducted on staticallycompacted fine-grained soil (i.e., compacted Indian Head till)in a laboratory environment to study the validity of themodified  and  methods. The test results of the study
presented in this paper show significant increase in shaft
capacity due to the contribution of matric suction. The
modified ,  and  methods provided reasonably goodcomparison with the model pile load test results conducted ina laboratory environment.
The authors based on the experience of the present study
suggest different techniques of testing to alleviate some

8 REFERENCES
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[2]

[3]
[4]
[5]
[6]

[7]

[8]

[9]

37

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Mechanics, 2007, pp. 483-494.
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capacity of unsaturated fine-grained soils,” in Proc. 62nd CanadianGeotechnical Conf & 10th Joint CGS/IAH-CNC Groundwater Conf.,
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[10] Vanapalli SK, Oh WT, “Interpretation of the bearing capacity of an
unsaturated soils extending the effective and total stress approaches,” inProc. 5th Int. Conf. on UNSAT, 2010, pp. 1223-1229.
[11] Douthitt B, Houston W, Houston S, Walsh K, “Effect of wetting on pile
friction,” in Proc. of 2nd Int. Conf. on UNSAT, 1998, pp. 219-224.
[12] Costa YD, Cintra JC, Zornberg JG, “Influence of matric suction on the
results of plate load tests performed on a lateritic soil deposit,”
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[13] Georgiadis K, Potts DM, Zdravkovic L., “The influence of partial soil
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[14] Vanapalli SK, Eigenbrod KD, Taylan ZN, Catana C, Oh WT, Garven E,“A technique for estimating the shaft resistance of test piles inunsaturated soils,” in Proc. 5th Int. Conf. on UNSAT, 2010, pp.
1209-1216.
[15] McClelland B, “Design of deep penetration piles for ocean structures,” J.
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[16] Tomlinson MJ, “The adhesion of piles driven in clay soils,” in Proc. 4th
Int. Conf. on Soil Mech. and Foundation Eng., vol. 2, 1957, pp. 66-71.
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[19] Fredlund DG, Xing A, Fredlund MD, Barbour SL, “The relationship of
the unsaturated soil shear strength to the soil-water characteristic
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[20] Vanapalli SK, Fredlund DG, “Comparison of different procedures topredict unsaturated soil shear strength,” in Proc. the Conf. on Advancesin Unsaturated Geotechnics Geo-Denver, 2000, 195-209.
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177-184.
[23] Kerisel J, “Vertical and horizontal bearing capacity of deep foundationin clay,” in Proc. Symp. on Bearing Capacity of Settlement ofFoundations, 1965, pp. 45-52.
[24] Sowers GB, Sowers GG, Introductory soil mechanics and foundations.
New York: The Macmillan Company, 1970.
[25] API Recommended practice for planning, designing and constructing
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[26] McCarthy DR, Essentials of soil mechanics and foundations. Reston
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[31] Vanapalli SK, Taylan ZN, “Modeling the load carrying capacity ofsingle piles in unsaturated soils using the modified  and the 
methods,” in Proc. 13th Int. Conf. on IACMAG, 2011, pp. 599-607.
[32] Vanapalli SK, Taylan ZN, “Model piles behavior in a compacted
fine-grained unsaturated soil,” in Proc. 15th European Conf. on
ESMGE, 2011, pp. 683-689.
[33] Gan, JKM, Fredlund DG, Rahardjo H, “Determination of the shear
strength parameters of an unsaturated soil using the direct shear test,”
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[34] Vanapalli SK, Oh WT, Puppala AJ, “Determination of the bearing
capacity of unsaturated soils under undrained loading conditions,” inProc. 60th Canadian Geotechnical Conf., 2007, pp. 1002-1009[35] Vanapalli SK, Taylan ZN, “Estimation of the shaft capacity of model
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38

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Problematic Soil: In Search for Solution
Bujang B.K. Huat
Department of Civil Engineering, Universiti Putra Malaysia, Serdang, Selangor, Malaysia



ABSTRACT: Soils are considered to include all naturally
occurring loose or soft deposit overlying the solid bedrock. They areformed of the disintegration and decomposition of rocks and also bydecomposition of organic materials. Compared with rocks, soils aresofter in term of strength and more compressible, thus giving more
problems to engineering compared with rocks. But is soil really aproblem, or the theory is? Perhaps the later is more valid.
In Malaysia, soils found can generally be termed as tropical soils.
They range from residual soils which normally are soils formed byin-situ weathering of parent rocks to soft clays and organic soils
including peat. Soft soils generally have low shear strength and highcompressibility. As such, problems like large settlements, flooding
and damage to infrastructure due to soil subsidence, foundation
failure due to lateral movement of soil, etc, are common. Residuals
soil on the other hand are generally considered to be better in term ofengineering properties compared with the soft soils, but by virtue of
their existence in sloping ground, landslides are a major concern.
Some 400 landslides have been reported (more than 30 were major
landslides), involving both cut and natural slopes with a total loss ofmore than 200 lives and billion of Ringgit in damage to properties
have been reported in the country over the last two decades or so.
In meeting the above mentioned challenges, some research works
have been done at Universiti Putra Malaysia. These include proper
analysis and design of piles against lateral loading; fundamental
understanding on the properties and behavior of soft soils, both soft
clays and peat; methods to improve/stabilize the soils; slope
assessment system for landslide prediction, bio-engineering
technique to stabilize slope; and use of waste materials as an
economical solution to repair slopes.
In this paper, some recent works in particular those related to peat,
will be described.

Keywords: Bio-engineering, Landslide, Soft Soil, Malaysia,
Soil Stabilization(igneous, sedimentary or metamorphic), mainly chemical
weathering, is the main agent for soil formations in thetropical country like Malaysia. The soils formed by theweathering are largely left in place, thereby literally called
residual soil, and whose character depends on the parent rockit develops from. In addition to the residual soils, transported
soils are also found in Malaysia. Another type of soils that canbe found in Malaysia is the organic soils. Peat actually
represents an accumulation of disintegrated plant remains,which have been preserved under condition of incomplete
aeration and high water content [1-6].
2. PROBLEMS ASSOCIATED WITH SOFT ANDSLOPING GROUND
Many towns and cities are actually located close to rivermouths, deltas, lakes or alluvial flats. These are usually theareas of soft compressible soils. The soils usually have low
undrained shear strength with high void ratio (highcompressibility). Other areas that bring many problems toengineering are the swampy and peaty area where the soils
have high organic content.
Landslides, problems usually associated with slopes(sloping ground), are serious geologic hazard common mostcountries of the world. Since 1975, some 400 landslides has
been reported in Malaysia (more than 30 were major
landslides), involving both cut and natural slopes with a total
loss of more than 200 lives and billion of Ringgit in damage
to properties[1-6].

3. SOME RECENT RESEARCH WORKS DONE AT
UNIVERSITI PUTRA MALAYSIA
1. INTRODUCTION

3.1.1 Laboratory Simulation of Pile In An Excavation
For this study a model test tank of 300 mm (W) x 800 mm(L) x 600 mm was used. The model test pile was instrumented
with strain gauges. The soil was well graded medium finesand with friction angle of 31o and the max sand density was
13.8 kN/m3.
Figure 1a shows the initial laboratory setup. Figure 1b
shows the FEM model in PLAXIS 3D. Figure 1c and 1dshows the results of the FEM model by PLAXIS 3DFoundation. Both the experimental and FE results wereapparently in good agreement [7-9].

Malaysia lies in areas with tropical climates (the A
category climate in the Köppen Classification System). This
climate is extensive, occupying almost all of the continents
between latitudes 20°N to 20°S of the equator.
The key criterion for an A category climate is for thecoolest month to have a temperature of more than 18 ºC
making it the only true winterless climate category of theworld. Another characteristic is the prevalence of moisture.
Warm, moist and unstable air masses frequent the oceans atthese latitudes. As a consequence, this climate zone hasabundant sources of moisture giving rise to high humidity.
Due to this climatic condition, weathering of parent rocks

39

40

condition. The width of the model was three times of pile
diameter. Few of the parameters were based on the pileproperties, soil properties and excavation configuration.
Figure 3 shows the model in 3D FE stimulation.

1000

0 day

900

moist cured

800

moist cured + 10 kPa
surcharge load
air curedU C S (kPa)

700
600
500
400
300
200
100
0

Undisturbed
fibrous peat

5% OPC

15% OPC

30% OPC

50% OPC(a) Unconfined Compressive Strength of untreated fibrous peat andpeat treated with OPC after 90 days curing
40

unsoaked

Figure 3. Total horizontal soil displacement in FE modelsoakedCBR (%)

30

The soil stiffness ranges from very soft to medium stiff clay
underlain by a hard layer. The effects of few parameters areclearly shown in design charts namely the soil stiffness, spun
pile diameter and depth function. These few parameters
showed significant effect when applied to cases with different
soil stiffness (cu=10 kPa, 20 kPa and 30 kPa), pile diameter(300 mm, 450 mm and 600 mm) and depth function ranging
from 2 to 7. Figure 4 shows the design charts for a particularconfiguration [7-9].

20
10

Fp

F

F

5%

PP

C
O
P
0%

t+
5

O
P
0%

t+
5
Fp

t+
3

t+
1
Fp

C+
0.
1

C

PP
5%

C+
0.
1

0%

O
P

Fp

F

O
P

PP
5%

0%
t+
3

Fp

C+
0.
1

PP

5%
t+
1

Fp

O
P
5%

%
t+
5
Fp

O
PC

O
PC

%

%

O
PC
+

0.
15

pe
at
s

t+
5
Fp

br
ou
.fi
nd
is
U

F

0

(b) CBR of untreated fibrous peat and peat treated with OPC andpolypropylene fibers after 90 days curing

CBR (%)

100

unsoaked - 24 hours

unsoaked - 28 days

unsoaked - 3 months

soaked - 3 months

75
50
25
0
PF
S tF
S tF
S tF
%P
+2%
+4%
4%
.15
C+
PF
PF
+0
OP
%P
%P
PC
5
5
%
O
5
0 .1
0 .1
5%
t +1
C+
C+
FP
t+1
OP
OP
FP
5%
5%
t +1
t +1
FP
FP

O
5%
t +1
FP

PC

(c) CBR of untreated fibrous peat and peat treated with OPC,
polypropylene and steel fibers after 3 months curing

Figure 4 Maximum moment for pile in excavation120

unsoaked - 24 hrs

unsoaked - 28 days

unsoaked - 3 months

soaked - 3 months

100
CBR (% )3.2 Stabilization of Peat
3.2.1 Shallow Stabilization

80
60
40
20

The effects of adding various percentages of chemicalbinders (OPC) from 5 to 50% in terms of wet weight of peatand other additives - polypropylene fibers, steel fibers, silica
fume (micro silica), ground granulated blast furnace slag andfly ash on the strength of fibrous peat were investigated.

0

t +2
FP

O
.5%

C
C
C
A
PC
FA
FA
FA
OP
OP
OP
%F
%O
7 .5
.5%
1 5%
5%
0%
0%
+2 .5
C+
t +5
C+
t +1
t +5
t+3
+ 22
FP
PC
OP
OP
FP
FP
FP
PC
.5%
%O
15%
+
t +7
.5
t
2
P
F
FP
t +2
FP

(d) CBR of untreated fibrous peat and peat treated with OPC and fuelash after 3 months curing

The results were shown in Figure 5. In general adding thechemical additives proved to have increased the soil strength
substantially. Additions of additives are also beneficial,
except for blast furnace slag, and fly ash where there appear
to be no additional gain. Air curing was found to be generally
better than moist curing. The best additive was apparently thepolypropylene fibers (about 0.15% or 1.5 kg polypropylene
fiber per cubic meter of peat at its natural water content)[10-11].

Figure 5. Effects of OPC and additives on peat.
3.2.2 Deep Stabilization
Deep stabilization involved installing ‘columns’ of peat
added with OPC and other additives into pre-bored hole set inpeat sample. This stabilization technique is akin to those of
cement or sand column as applied to soft clays. Peat withcement column was then tested in triaxial for strength andRowe cell for compressibility.

41

The cement column technique involved casting in place a
20 mm diameter column (which gave an area ratio of 0.16)
inside a 50 mm sample. The cement column was formed byadding peat with varying percentage of cement (OPC), theproportions of cement to peat ratio chosen are (100:0),(75:25) and (25:75) compacted to inside a PVC tube. The
column was cured for 45 days in a soaking basin, before being
inserted inside a pre-bored hole.
Figure 6 illustrates the effect of the cement column on theshear strength and compressibility of peat, measured withtriaxial and Rowe cell test respectively. It was found that thecement column significantly improved both the shear strength(100% increase) and compressibility of peat, in particularcolumn with higher dosage of cement. The peat gained
strength with the addition of cement and probably thepresence of cations initiated the flocculation because of thecementation phenomena among the soil particles and cement.

Figure 7 Test tank
Precast cement column, 1.0m long and 200 mm and 300
mm diameter were made of peat mixed with OPC andadditives (polypropylene fiber, silica fume, polypropylene
and steel fibers). The cement column was again shown to bebeneficial in improving the load carrying capacity of peat
(Figure 8)[10-11].
Load (kN)
0

2

4

6

8

10

12

14

16

18

20

22

0

Deformation (mm)

20
40
60
80
100
120
Untre
FPeat FPt+15
%OPC

140

(a) Shear strength parameters of peat160

FPt+15%
OPC+SFU

FPt+20%OPC

FPt+15%OPC+PPF

FPt+15%OPC+
PPF+StF

FPt+15%OPC(300 m mdia.)(a) Load-deformation of untreated peat and peat treated withprecast cement column
0

0.5

1

Load (kN)

1.5

2

2.5

0

Deformation (mm)

20
40
60

loading
unloading80100(b) Compression index of peat120

plain fibrous peat

FPt+15%OPC(300 mm dia. column)(b) loading-unloading

Figure 8 Load carrying capacity of fibrous peat with andwithout precast cement column
3.2.3 Stabilization by Injection and Vacuum
The Deep Mixing Method (DMM) is today accepted
world-wide as a soil improvement method and is based on
mixing binders, such as cement, lime, fly ash and otheradditives, with the soil to form columns of a hardening
material. DMM coupled with vacuum technique appears to be
the solution for deep stabilization of peat. To do this, alaboratory scale of new DMM technique (injection-vacuum
apparatus) was specially designed and developed as shown in
Figure 9. In this study, the effects of cement, sodium silicate,calcium chloride, and kaolinite grouts on peat properties in
new DMM technique were investigated [12-14].

(c) Secondary compression (re-compression) index

Figure 6 Effects of precast cement columns on peat
To simulate a field condition, test at larger scale was carriedout (Figure 7). Instrumentation included pore pressure gauge,
LVDT, horizontal strain gauge and pressure cell.

42

The results showed that the application of vacuum would
help speed up the stabilization process. The low strength andloose matrix of peat meant that usage high power injection
pump for stabilizing is not possible. In this method low power
injection pump is applied by aiding of vacuum pump toincrease the stabilized area. In comparison with conventional
vacuum and preloading method as well as new injection
methods, this method is believed to be morecost-effective.[12-14].

Figure 11 Electro-migration test system
The results showed that the coefficient of electro-osmotic
conductivity and the 3-day energy expenditure was found toincrease with increase in the degree of peat humification(Figure 12). The H8 peat had the highest electro-osmotic
conductivity and energy expenditure. Peat had a net negative
charge and the direction of electro-osmotic flow was fromanode to cathode. The study showed all charges on the soils’
surfaces were strongly pH-dependent. The very highly
decomposed (H8) peat had the significant differences inelectro-osmotic properties in comparison with H5 and H3peat [15-17].

Figure 9 Small scale of new DMM technique3.2.4 Electro-osmotic properties of peat
We began a study in 2007 to look at the role of organic
matter on electro-osmotic properties of peat; and toinvestigate the behavior of electro-osmotic flow in peat. The ζ
was measured with a zeta-meter as a function of pH values
ranging from 1.91 to 11.5. All measurements were made in
0.0001 M NaCl solutions and pH adjustments were madeusing dilute HCl or NaOH solutions.
Figure 10 shows the relationship between peat zeta
potential (ζ) of peaty soils with soil pH. The variations in ζwith pH were probably related to the nature of electricalenergy field in peaty soils. [15-17](a) Coefficient of electro-osmotic conductivity-degree of
humification-energy expenditure(a) Zeta potential – pH

Figure 12 Electro-osmotic phenomena in peat

Figure 10 Zeta potential – pH relationship3.2.5 Electro-osmotic Phenomena in Peat
The test apparatus consisted of an acrylic unit with a central
cylinder of 150 mm in length and 169 mm in internal diameter,
as shown in Figure 11. The volume of both the cathode andthe anode compartments were 2243 mL. Titanium disks wereused as the electrodes.
Peat samples were treated for 3-days period. The effluent
was collected to calculate the coefficient of electro-osmotic
conductivity.

43

3.3 Slope Assessment System for predicting landslide
In our work, we developed a slope assessment system(SAS) for cut slopes underlain by granitic formation using
discriminant analysis. From the available data, the following
physical parameters were selected, namely: slope feature,
location, height of slope, angle of slope, feature aspect indegrees, plan profile, cross profile, distance to ridge, batter,slope shape, and main cover type. These data in form ofcontinuous variables/parameters were transformed into
various classes and were used in the statistical analysis andregression equation for the computation of instability score(individual discriminant scores). The database was compiled
between 1994 and 2003 on a total of 36 slopes, 21 slopes fromthe Tapah – Cameron Highland road, and 15 slopes from
Kuala Lumpur – Bentung old road [18-21].
From the results, it was seen that the higher the significant
value of F to add new and to delete variables, higher the
Eigen-values and χ2 produced. The Wilk’s  however became
smaller. Generally the number of significant parameters
would increase with increase of the significant value of F to
add and to delete. Discriminant function of the 86 failed and
53 not failed slopes was computed. The boundary of
discriminant function separating these two groups (failed andnot failed slope) was calculated using the average of the twogroup’s mean. Groups mean for not failed and failed slopes
were -0.91 and 0.58 respectively. The value of discriminantfunction separating these two groups (noted as g) could be
calculated. g = (Yf + Ys) / 2, where, Yf = Mean of failed group;
Ys = Mean of not failed group.Value of g for the model is-0.165.Using this g value, the boundary condition separating
failed and not failed slopes is as follow; Not failed if Y <-0.165, otherwise failed. The hazard rating was designed
using the maximum and minimum value of discriminantfunction. The maximum value of discriminant function was
5.906 and minimum value was -7.083. Table 1 below shows
the hazard rating designed for the SAS [18-21].

Our first challenge was to identify suitable plant species forsuch an application. Our study led us to 10 potential tropicalplant species. Based on screening trial and observations of
distribution/location and shape of growing roots three species
were initially identified as the best candidate for the live pole,
namely: Hibiscus tiliaceus (Ht), Dillenia indica (Di) andDillenia suffruticosa (Ds). The shade house experiment was
replicated but with mineral soils (aerisols, ferralsols, histosols,
luvisols, regosols, gleysols, fluvisols) that commonly formed
the Malaysian slopes. From this experiment, only 2 species,
Ht and Ds were found later to be the most suitable (Figure13)[22-25].

Hibiscus tiliaceus

Dillenia indica

Dillenia suffruticosa

Figure 13 Rooting trial (root growth along stem)
We carried out shear test to measure the effect of root onsoil strength. Unfortunately the conventional shear box was
not good enough to represent the effect of root in a soil mass.
For this, we have to design and fabricate our own large shear
box. This box had a sample size of 300mm x 300mm x
200mm, compared with conventional equipment whose
sample size is only 50x50mm.
The results showed that the presence of roots had
significantly improved the shear strength of the soil and it
also showed that the effect was mainly on the cohesion. Rootof Ht has enhanced the cohesion component of shear strengthby 593% (at 30 cm depth) and 722% (at 50 cm depth) as
compared to the unplanted soil.

Table 1 Hazard rating designed for the new SAS
Range
Rating2.871 to5.906
Very high-0.165 to
2.871
High
-3.459 to-0.165Low-7.083 to-3.459Very low
The results of the comparative study are as shown in Table 2.

To assess the suitability of the plant species and live pole inreal situation, a field trial was carried out. A meta-stable slope
in the vicinity of the university (UPM) campus was selected.
The planting operation was done in two days. All fresh cuttingsof Ht and Ds had initial lengths of 2.10 to 2.30, diameters
between 50 to 70 mm at upper end and 50 to 80 mm at butt end;
and almost straight, smoothly tapered and with no bends or
branch points forming large bifurcations. At the end of the 12months monitoring period, 2 Ht and 2 Ds live poles were
exhumed in order to study the root growth and properties ofthe live poles in a field condition as shown in Figure 14.
A visual comparative study of Ds and Ht live poles showedthat about 25% of the embedded length of Ht was rooted, forDs it was about 31% Figure 15). However, the roots of Ht
seemed longer (230 to 1230 mm long) and thicker (0.5 to 5.7
mm diameter) compared with roots of Ds (length: 350 to 1060
mm; diameter: 0.5 to 2.5 mm).

Table 2 Evaluation SAS in predicting landslides(1) Number of assessed slopes
36(2) Numbers of actual landslide or25
failed slope(3) Numbers of Slope classified as
28high hazard(4) Number of slopes classified as24high hazard that actually failed(5) % of correctly classified failed
96%
slopesThis SAS appears to be satisfactory.
3.4 Use of Bio-engineering for Stabilizing Slope
Bio-engineering in our context refers to the use of plants toenhance the stability of hill slopes primarily against
shallow-seated. This is the basic principle of what is termed
as the live pole technique.
44

Next we needed to design and study the performance of the
propose tire system. A full scale field trial was carried. A totalof 2100 numbers of scrap car tires in 25 layers was used. It
only took 5 unskilled workers 20 workings days to complete
the structure. Instrumentation like settlement plates andpressure cells were installed to monitor the performance ofthe trial wall (Figure 16).

Figure 14 Exhumed live poles from the site

Figure 16 Constructing the trial scrap tire wall(a) Ds pole

The following software (GawaWin, Prokon and MacStars)
were used for the preliminary design. Some basic design
assumptions were (1) maximum bearing capacity of base was
100 kPa, (2) effective internal friction was 28o, (3) no waterlevel, (4) surcharge was 10 kPa, (5) factors of safety against
sliding, overturning, bearing failure and overall slope stability
were 1.5, 1.5, 3.0 & 1.5 respectively. As shown in Figure 17,
the propose scrap tire system was apparently the most costeffective for wall of 2m to 6m. The significant advantage ofthe propose scrap tire system was the reduced cost of wallmaterials and construction. Most of the wall components(scrap tires and back fill soils) were recycled material [26].

(b) Ht pole

Figure 15 Roots of exhumed Ht and Ds pole
A finite element analysis was carried out to illustrate thebenefit of live pole in stabilizing slope against shallow
failures using 2D Plaxis. As expected, the factor of safety ofthe slope increased markedly. The live pole reinforcement
was apparently more effective if poles were installed in closer
array and soil had higher original cohesion [22-25].

Cost Comparison of Various Retaining Walls

Wall Cost (per m, Ringgit Malaysia)

6000

3.5 Low Cost Repair for Failed Slopes
In order to apply scrap car tires as slope repair, we began bydesigning suitable tests and then tested the locally produced(scrap) car tires. We carried out tensile tests on commonly
available sizes, i.e. R12 to R15. Mean tensile strength of
55.81 kN, with standard deviation 15.19 kN was obtained
from the pull-out test, with statistic probability of tensile
strength greater than 20 kN was 99%. Two attachment
systems were considered and tested in laboratory for tension,
the wire rope and U-clip, and polymer rope of different
numbers of wrap and knots. For the wire rope and U-clip
system, a 12 mm diameter wire rope with its compatible size
U-clip was selected for testing purpose. Whilst for polymer
rope, 12 mm diameter polypropylene was chosen for ease ofhandling and economy. The rope was tightened on the crossbeam of tensile machine and pull until the rope rupture. The
polypropylene rope had a working load of 25kN was a
suitable attachment for scrap tire wall.

5000
Tire Wall

4000

Terramesh
RE Wall

3000

Stone Pitching
Gabion2000

RC Wall

1000
0
2

3

4

5

Wall Height (m)Figure 17 Cost per linear meter of various retaining wall
systems versus height
4 CURRENT RESEARCH
We are currently working on some research projects
including:
4.1 Electro-Biogrouting of Soft Soils
Biogrouting is a new method to stabilize sandy soils
through precipitation of CaCO3 between grains. This process
decreases the permeability about 98% and increases the soil
45

[5]

Huat, BBK, Han KK, Loh, WC & Aziz A, Retaining Structures: Design
and Construction in Tropical Soils. Universiti Putra Malaysia Press,
2006.
[6] Huat BBK, Organic and Peat Soils Engineering, Serdang: University
Putra Malaysia Press, 146 p.
[7] Kok ST & Huat BBK, Numerical Modeling of Laterally Loaded Piles.
American Journal of Applied Sciences. Vol. 5(10), 2008,pp. 1403-1408[8] Kok ST, Bujang BKH, Jamaloddin N, Mohd Saleh J, Gue SS, A
Review of Basic Soil Constitutive Models for GeotechnicalApplication, EJGE Vol 14, 2009, pp 1-18.
[9] Kok ST, Bujang BKH, Jamaloddin N, Mohd. SJ, Gue SS, A Case Studyof Passive Piles Failure in Open Excavation. Journal of Deep
Foundation Institute, Vol. 3, No. 2, 2009, pp. 50-57.
[10] Kalantari B, Prasad A. & Huat BBK, Stabilising Peat Soil with Cement
and Silica Fume. Proceedings of the Institution of Civil Engineers:
Geotechnical Engineering. 163 GE1, 2011, pp. 33-40.
[11] Kalantari B & Huat BBK, Improving Unconfined Compressive
Strength of Peat with Cement, Polypropylene Fibers, and Air Curing
Technique. Global Journal of Researches in Engineering. Vol. 10(1)2010, pp 9-15.
[12] Huat, BBK, Kazemian,S & Barghchi M, A State of Art Review of Peat:
General Perspective. International Journal of the Physical Sciences 6(8),2011, pp. 1988-1996.
[13] Kazemian, S, Huat BBK, Prasad A & Barghci M, A State of Art
Review of Peat: Geotechnical Engineering Perspective. International
Journal of the Physical Sciences, Vol. 6(8), 2011, pp. 1974-1981.
[14] Kazemian S, Huat BBK, Effects of Aggressive pH Media on Peat
Treated by Cement and Sodium Silicate Grout. J. Cent South Univ
Technol. Vol. 18, 2011, pp 840-847[15] Asadi A, Huat BBK, Hanafi MM, Mohamed, TA & Shariatmadari N,
Chemico-geomechanical Sensitivities of Tropical Peat to Pore Fluid pH
Related to Controlling Electrokinetic Environment. Journal of theChinese Institute of Engineers. Vol.34(4), 2011, pp. 481-487[16] Asadi A, Huat BBK, Hanafi MM, Mohamed TA, and Shariatmadari N,
Role of Organic matter on Electroosmotic Properties and Ionic
Modification of Organic Soils. Geosciences Journal, Vol 13(2) 2009, p
p 175-181.
[17] Asadi A, Huat BBK, Hanafi MM, Mohamed TA, Shariatmadari N,
Physicochemical Sensitivities of Tropical Peat to Electrokinetic
Environment. Geosciences Journal, Vol 14(1), 2010, pp. 65-75.
[18] Singh H, Huat, BBK & Jamaludin S, Slope Assessment Systems: A
Review and Evaluation of Current Techniques Used for Cut Slopes in
the Mountainous Terrain of West Malaysia. Electronic Journal of
Geotechnical Engineering, Vol.13, 2008, pp. 1-24.
[19] Jamaludin S, Huat BBK & Omar H. Evaluation and Development of
Cut-Slope Assessment Systems for Peninsular Malaysia in Predicting
Landslides in Granitic Formations. Journal Technology B. Universiti
Technology Malaysia. Vol. 44B, 2006, pp 31-46.
[20] Jamaludin S, Huat BBK and Omar H. Evaluation of Systems for
Predicting Landslides of Cut Slopes in Granitic and Meta-sedimentFormations. American Journal of Environmental Sciences. Special
Issues on Recent Advances in Landslide Engineering and Technology.
NY, USA. 2(4), 2006, pp. 135-141[21] Huat BBK & Jamaludin S, Evaluation of Slope Assessment System in
Predicting Landslides along Roads Underlain by Granitic Formation.
American Journal of Environmental Sciences. Vol. 1(2), 2005, pp.
90-96[22] Huat BBK & Kazemian S. Study of Root Theories in Green TropicalSlope Stability. Electronic Journal of Geotechnical Engineering.
Vol.15Q, 2010, pp. 1815-1824.
[23] Mafian S, Huat BBK, A Rahman N, Singh H. Potential Plant Species
for Live Pole Application in Tropical Environment. American Journalof Environmental Sciences, Vol. 5(6), 2009, pp. 759-764.
[24] Mafian S, Huat BBK.& Ghiasi V. Evaluation of Root Theories andRoot Strength Properties in Slope Stability. European Journal of
Scientific Research. Vol.30(4), 2009, pp. 594-607.
[25] Mafian S, Huat BBK, Barker DH, Rahman NA & Singh H, Live Polesfor Slope Stabilization in the Tropical Environment. Electronic Journalof Geotechnical Engineering. Vol.14G, 2009, pp. 1-25.
[26] Huat BBK, Aziz AA & Loh WC, Application of Scrap Tires as EarthReinforcement for Repair of Tropical Residual Soil Slope. Electronic
Journal of Geotechnical Engineering. Vol. 13 B, 2008, pp. 1-9.

strength up to 12Mpa. It is created by bacterium Bacillus
Pasteurii and enzyme urease that hydrolyze urea to carbonicdioxide and ammonia. A major problem in biogrouting is the
distribution of bacteria injected into the soils. We are going touse electrokinetics in order to transport bacteria towards thecathodes to get a uniform distribution of bacteria and enzyme
in the soft soil mass.
4.2 Acid Rain Intrusion Effects on Slope Failure
Phenomena and Mechanisms
Acidic deposit that fall to Earth from the atmosphere isnamed acid rain. It is caused by sulfur dioxide (SO2) andnitrous oxide (N2O). Acidic rainfall as an environmental
factor has a potential to affect slope stability. However, theslope failure phenomena and mechanism cannot be explainedwith a simple mechanical-physical concept. A new approachis needed to explain the underlying mechanism of waterattack on soil using physicochemical concept including ion
exchange effects.
This fundamental research will integrate and blend
traditional theory with physicochemical concepts in order toprovide a framework for the analysis of soil behavior underacid rain intrusion effects. The obtained results will be
discussed in the light of chemico-mechanical model. The
outcome will explain the “why” and “how” the mechanicalsoil behavior under acid rain intrusion effects in anenvironmental context.
5 CONCLUSIONSIn this paper, we presented some research works which havebeen carried out at Universiti Putra Malaysia. These include
proper analysis and design of piles against lateral loading;fundamental understanding on the properties and behavior ofsoft soils, both soft clays and peat; methods toimprove/stabilize the soils; slope assessment system forlandslide prediction, bio-engineering technique to stabilize
slope; and use of waste materials as an economical solution to
repair slopes.
6 ACKNOWLEDGMENT
The financial assistance from the Ministry of Science,
Technology and Innovation Malaysia, as well as the
Universiti Putra Malaysia for the researches mentioned in thispaper is gratefully acknowledged. So are the input of myresearch students, past and present.
7 REFERENCES
[1]

[2]
[3]
[4]

Huat BBK., Ali FH, Barker D, Singh H. & Omar H., Landslides in
Malaysia:
Occurrences,
Assessment,
Analysis
andremediation/Preventive Solutions. Universiti Putra Malaysia Press.
2008, 468p.
Huat BBK & Ali F, Essential Soil Mechanics for Engineers. Universiti
Putra Malaysia Press,2007, 518p.
Huat, BBK & Ali F, Ground Improvement Techniques. Universiti Putra
Malaysia Press, 2007, 383p.
Huat BBK, Ali F, Omar H & Singh H, Foundation Engineering:
Design and Construction in Tropical Soil. Taylor and Francis, United
Kingdom, 2006, 243p.

46

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Characterization and Modeling of Various Aspects of Pre-failure Deformationof Clayey Geomaterials – Fundamental Theories and Analyses
John N. MUKABI., EngineerIntel – Research & Development Department, Kensetsu Kaihatsu Consultants,
Nairobi, Kenya

the use of sophisticated methods, determines relatively
precise and reliable Value Engineering (VE) based
cost-effective design parameters taking into serious
account prevalent loading factors, environmental
conditions and structural sustainability.

ABSTRACT: The recently preferred Performance–Based
Design (PBD) fundamentally entails that, deformation in groundand foundation soils, along with the reciprocal and/or
retrospective structural deformation and stress states, arecomprehensively analyzed by adopting sophisticated methods,
particularly for structures within high exposure seismic action
regions. The soil mechanics concept of deformation resistance,
which principally constitutes the fundamental postulates onkinematic hardening, is considered as one of the most vital
conceptual approach in this regard.
In this part of the paper therefore, fundamental theories related to
the recently proven Modified Critical State Theory (MCST),“Consolidation and Shear Stress Ratio” (CSSR) functions as wellas the ) mathematical concepts on Yield Strain Limit (YSL
proposed in this Study, that may enable the effective use ofretrospective analysis in simulating, predicting andcharacterizing the variation in the behavior of clayey composite
ground and geomaterials under various loading conditions, areanalytically introduced and subsequently applied.

1.2 Methods of Testing and Geotechnical Investigation
Laboratory tests were comprehensively performed inJapan by using sophisticated automated triaxial testing
systems equipped with a local strain measurement device;
the Local Deformation Transducer (LDT), whilst various
geophysical methods of survey were employed inundertaking the in-situ geotechnical investigations. The
seismic survey was mainly adopted in determining thein-situ initial elastic and/or shear moduli. On the other
hand, the Transient Electromagnetic Method (TEM)
utilizing sounding technology was used in deriving in-situ
bearing capacity, strength and elastic stiffness bycorrelating the measured resistivity to the geotechnicalparameters through a newly proposed quasi-empiricalmethod [1]. The main mechanical methods adopted werethe Standard Penetration Test (SPT) and the Dynamic
Cone Penetration Test (DCPT). Standard and modified
laboratory tests were also carried out extensively in the
East and Central African Region using conventional
triaxial and unconfined compression testing equipment.
Details of the methods of testing and testing equipment canbe referenced from [2], [3].

Keywords: deformation, yield sub-surfaces, consolidationstress-strain history, loading rate, reconstitution
1. INTRODUCTION
1.1 Imposed Geo-changes in Ground and Geomaterial
When subjected to various modes of loading andenvironmental conditions, most ground and geomaterials
undergo geotechnical changes that are, as far as thegeotechnical engineer is concerned, of great exigency interms of project costs, particularly in developing countries
which suffer, more predominantly, in lack of adequate
project funding. On the other hand, although conventional
approaches such as the limit equilibrium-based methods
provide information to resist a design seismic force forexample, they are not able to predict structural
performance under critical loading conditions.
Furthermore, notwithstanding the simplicity andpopularity of the Equivalent Linear procedure, its main
limitation is characterized by the lack of providing any
prediction of earthquake induced permanent displacements.
Due to these factors, and the increased frequency, intensity
and changes in trend of seismic action in recent times, andin consideration of the degree of devastation caused withinshort time periods on civil engineering structures that havebeen subjected to earthquake and tsunami loading, it isvirtually a definitive prerequisite that multi-stage
advanced and sophisticated methods of analyses with theeffective ability to probe and/or predict the transformed
ground and material geo-characteristics, be employed.
It is therefore imperative that, for structures which require
higher performance, the geotechnical engineer, through1.3 Geomaterials Adopted for Laboratory Testing
The well cemented and highly structured Pleistocene clays
adopted for this study were sourced and tested in Japan.
The clays were mainly sourced from: Osaka Bay, (theconstruction site of the Kansai International Airport);
Tokyo Bay Clay (the construction site of the Trans Tokyo
Bay Highway connecting Kawasaki City and Kisarazu
City in Chiba Prefecture); OAP Clay (the construction siteof skyscraper buildings in Osaka City), and Suginami Clay(the construction site of the Kandagawa Flood Control
Project in Tokyo). Other clays that were investigatedinclude the Ariake Holocene Clay, which is a very soft,sensitive and highly structured clay underlying the Ariake
area in Kyushu, Japan [2].
Lateritic gravel and black cotton soil, which are in
predominant existence in most parts of Africa, weresourced from various construction sites in East and Central
Africa, and tested in the laboratory under varying modes ofloading and geotechnical conditions. Black cotton soils,
formed in the Quaternary period from in-situ weathering ofbasic igneous, metamorphic and pyroclastic rocks, arehighly expansive, problematic and extremely susceptible
to moisture ~ suction variations.
47

2.

FUNDAMENTAL
THEORETICAL
CONCEPTUAL CONSIDERATIONS

AND

2.1 Brief Background of the MCS Theory
The fundamental postulates of the MCST (Modified
Critical State Theory) were first proposed by [4] while
investigating the effects of consolidation stress ratio andstrain rate on the peak stress ratio of clay. In that study,
they concluded that, possibly due to different structure
and the maximum stressformation, the stress rate at q
increased as the consolidation stress ratio
ratio q/p
σ /σ decreased. Following further research, and byK
demonstrating the inadequacies of the conventional
Critical State Theory (CST), [5], [6] proposed the need formodification of the CST and most of the theories based
thereon, which characterize the strength and deformation
behavior of clays assuming that failure for normallyconsolidated clays occurs at the Critical State Line (CSL)
irrespective of the drain condition, strain rate and stress
path traversed. The mathematical derivatives of the MCST
and its applications are reported in detail by [7]. Typicalexamples demonstrating that the CSL is not unique but
rather dependent on varying geo-changes and transformed
states are depicted in Figs. 2.1 and 2.2.
For clay subjected to cyclic loading at frequencies between
0.02Hz and 01.Hz, [8] defined a Cyclic Failure Line (CFL)
which is clearly strain rate dependent as shown in Fig. 2.1.
From this figure it can be noted that the gradient of thenormalized stresses reduces as the frequency increases.

Figure 2.2 Effective reconsolidation and shear stress paths
for undrained and drained Suginami clay specimens [2].
Consequently, under the MCST framework, the expressionthat locates the conventional CSL is therefore
modified to retrofit the modified CSLM as:
′ ′.

(1)

where,(2)
In this case, is the deviator stress, ′ is the stress pathlocating modifier, ′ is the mean effective stress, is the, which is defined as thetransformation function ofdifferential normalizing parameter of the angle of internal
friction,is the cubic exponent of the Consolidation
Stress Ratio (CSR) function, which is expressed as,.
On the other hand, the modified λM constant, which would
project the CSLM onto the υ−λℓ p' space when analyzingchanges in volume (voids) is expressed in the relation;(3)
hence,

Figure 2.1 Stress states at failure of clay under undrained
cyclic loading and Cyclic Loading’s Failure line (CFL) [8].
In their study on cemented sand subjected to undrained
cyclic loading, [9] also suggested that there exists a Phase
Transformation Line (PTL) which plots well above theconventional CSL. Their results also showed that the CFL
overshoots the CSL [(μf)CFL = 2.25, while (μf)CSL = 1.61].
On the other hand, Fig. 2.2 distinctly shows that the CSL is
not unique and varies with drainage conditions andconsolidation stress-strain history.
Consequently, the MCST takes into account threeimportant features: 1) the clay structure in relation to theconsolidation stress-strain-time history; 2) the initialpre-shear state of stress in relation to the modified CSL;
and, 3) the fact that under a constant rate of strain during
undrained shear, stress paths for higher consolidationstress ratio values traverse for longer periods and distances
to peak resulting in larger straining that contributes tolarger structural changes that culminates in the“overshooting” phenomenon of the conventional CSL [4].

′

(4)

′

(5)

and,

The deviator and mean effective stresses,
failure are then derived as:
′

and

/

at
(6)

and,
/
and
are material constants, while
where,
andare mapping constants.

(7),

2.2 Brief Introduction of CSSR Functions
The geo-mathematical derivations of the CSSR Functions
developed within the MCST framework and their48

For stiff geomaterials having relatively large initial andsecondary yield surfaces, the quasi-elastic properties areimportant as bench marks for both geotechnicalengineering analyses and the construction of a generalframework of pre-failure characterization.
Theoretically, as proposed by [11] and partly by [12], in
quantitatively determining the yield strains for therespective yield loci , the following considerations, most ofwhich are scientifically and quantitatively confirmed inthis Study for the first time, are made: 1) when the stresstranscends the YI and YS sub-yield boundary surfaces, thepreceding surface is kinematically engaged and draggedwith the stress point; 2) when YI boundary is engaged withsome associated energy loss, it is considered that particle
contacts start to experience local yielding due to normal
stress concentrations developing small plastic strains; 3)
significant plastic straining is delayed until the stress point
engages the YS sub-yield surface at which stage asignificant number of contact points probably start to failin shear, leading to inter-particle slippage resulting in amarked shift in the direction of the strain increment vector;4) as demonstrated in this Study, a further feature of the YS
envelope is that it defines the threshold at which drained
and undrained cyclic loading starts to affect the soil
significantly whereby sharp increases are observed in theratio of plastic to elastic strains, creep rates at constantstresses, energy dissipation in local cycles, hysteretic
damping ratio and rate of pore pressure development; 5)
the strains developed by a stress path which progresses
towards the YF locus (failure envelope) become
increasingly plastic and time-dependent in their nature(refer to Fig. 2.5); 6) a sharp change in stress path direction
leads to re-entry into the preceding yield surface and, 7)
destructuration leads to the disappearance of the existing
initial yield surface transferring the stress point to thesubsequent locus hence the prevalence of pseudo-elastic
characteristics depicting lower stiffness and/or apparently
longer pseudo-elastic range.
Furthermore, this Study also sets out to confirm, on thebasis of experimental data and analysis that, the size andshape of the yield surfaces, particularly YI and YS, aremostly influenced by: 1) recent stress and time history as
the stress point transcends to the current stress-state; 2)structuration due to ageing (Long Term Consolidation) andcementation occurring at the current stress-state as
particles agglomerate; 3) loading rate and path of
perturbing stresses or strains; 4) drainage conditions and,5) mode of loading static (monotonic, cyclic) or dynamic.
Based on the foregoing theoretical and practicalconsiderations, mathematical derivations which provide a
sufficiently applicable method of determining pre-failure
sub-yield strain limits and the respective elastic stiffnessand deformation modulus, are developed.
As shown in Fig. 2.4, an apparent (estimated) initialaYoung’s modulus E max
is first determined. An arbitrary

applications in correcting for the effects of SHANSEP(Stress History and Normalized Engineering Properties)
consolidation, ageing and reconstitution, are reported indetail by [10].
On the other hand, details of the pragmatic applications of
the versatile CSSR functions including correction for, andsimulation
of,
field
conditions,
retrospective
characterization and retracing of consolidation and shear
stress histories, multi-stage construction control offoundations and embankments on soft and problematicsoils and, in modification of constitutive models based on
the conventional CST, are discussed in detail by [7], [10].
2.3 Theoretical Considerations for Determination ofSub-yield Boundary Surfaces
Within the framework of classical visco-elasto-plastic
theory, the strength and yielding characteristics ofgeomaterials at relatively large strains are described bydefining a yield locus whereby the behavior within thislocus is considered either rigid, linear elastic or non-linear
elastic. Nevertheless, recent research has shown that such
simplifications are inadequate in characterizing the overall
stress-strain behavior of geomaterials over a wide range ofstress-strain. As a consequence, most researchers haveproposed multiple kinematic yield surfaces to better
describe the elasto-visco-plastic behavior observed within
the larger scale yield envelope.
Consequently, most of the recent constitutive models
describing the plastic behavior of geomaterials including
clays, sands and soft rocks subjected to cyclic loading arederived from the elasto-visco-plasticity theory along withthe concept of the non-linear kinematic hardening rule.
However, despite the various numerical and modelingresearch that has been reported, a quantitative method ofdetermining the limiting boundaries of the sub-yieldsurfaces based on experimental testing is not yet proposed.
In the method proposed in this Study, three kinematic
sub-yield surfaces, namely; YI: representing the Initial
Yield Surface within which the quasi-elastic properties aredefined, YS: representing the Secondary Yield Surface
which characterizes the intermediate non-linear andirreversible deformation behavior, and YT: the Tertiarybounding limit where large scale straining and yielding
occurs, are considered as schematically shown in Fig. 2.3.

modulus ratio, α is introduced within the region of small
strains whereby 1.0 < α ≦ 0.95. Considering theintersection properties of the parameters in Fig. 2.4, asquare relation expressed in (8) is developed.

Figure 2.3 A schematic representation of multiple
kinematic yield surfaces.

49





 E  E  YS   a    a S 2E   E   E a  E 
(13)  a    max   a    a  2E   E  E  Ea 0.1%
where,  Y  E     , , a  0.1%.
a a  a Y   a   





S

E

a
 a i   a o qi  qo  E max a a i   a o 

2

.

T

Scale Plastic Strain Limit)
This is the zone where large scale yielding occurs andmay be quantified from the proposed (14).
T  a  f   a T

are thence expressed as;a
E   dq d a    E max 2a a    a o 

and,



a
E   dq d a    E max
 2a  a    a o

 E   E  2E   E   a Y   a  f  (8)

The tangent moduli E  at point  and E  at point Ya E max
 E 
(14)  a      a    a  .
 2E   E  (strain at failure) and,
where, E  = modulus at(9) (10)

are arbitrarily designated deviator stresses atwhere, ,the corresponding , strain levels.
Solving (9) and (10) simultaneously and substituting for ain to (8), we obtain (11), which defines the initial Yield
Strain  a Y that locates the boundary of the quasi-linear a Y  E    a  .
TEAn example of the application of the proposed
mathematical relations for determining the sub‐yieldstrain limits is demonstrated in Figs. 2.5 and 2.6.

I

elastic range, i.e.; the Initial Yield Surface upontransformational mapping into the q ~ p  space.
Hypothetically, the magnitude is defined as the length ofthe major radial axis of the YI elliptical surface depicted inFig. 2.3.
(1)

 a Y : Tertiary Yield Strain (Pre-failure Large(3)

Figure 2.4 Graphical representation of stress-strain
parameters applied for mathematical determination of
Yield Strain Limits (YSL) [2].

 a Y : Initial Yield Strain (ELS)Ia
 E max
 E 
 a YI   a    2E   E   .

  a    a 

(11)



Figure 2.5 Yield surfaces traced from computed YS.

and,



 q  q    E  E 

i
o





2



a 
a  




2

       a ia o  100
E max 2 a i   a o 





(12)

where,

 a  
(2)

E
E

  a  , E    E  a 0.01% , E   E max .
a

 a Y : Secondary Yield Strain (Intermediate
S

Strain Limit)
The secondary yield strain which quantitatively defines theintermediate strain boundary is computed from (13).

Figure 2.6 Zoned sub-yield surfaces {(εa)ELS(YI)} ~{(εa)FYS(YF)} based on computed LTC/STC Yield Strains.

50

Strain {(εa)ELS(YI)} varies considerably with theconsolidation stress ratio and secondary consolidation
time history, the initial modulus is only dependant on the
consolidation stress ratio, whilst secondary consolidation
time history has virtually insignificant effects on thisparameter. From Fig. 3.1, however, it can further be noted
that, prior to the end of primary consolidation, the initialmodulus varies considerably with consolidation time.

The results of Long Term Consolidation (LTC) and ShortTerm Consolidation (STC) Osaka Bay clay specimens
shown in Figs. 2.5 and 2.6 clearly demonstrate that: 1)
multiple kinematic yield surfaces exist and, 2) the multiple
pre-failure sub-yield surfaces expand in size withprolonged ageing (LTC).
3 CHARACTERIZATION OF VARIOUS ASPECTSOF PRE-FAILURE DEFORMATION
3.1 Preamble
Adequate assessment, evaluation and analyses of soil
parameters to be adopted for design are crucial inenhancing the levels of precision of analyzing andcharacterizing the behavior of geotechnical structures.
Essentially, this implies that the choice and design of theexperimental technique to be adopted for soil
characterization is most vital since its effectiveness entails
a compromise between technical and financial factors. The
shortcomings however, of stress-strain models developedon the basis of laboratory test results of reconstituted
specimens are significantly demonstrated in this Study.
For example, [11] emphasized that realistic prediction ofground deformations and structural displacements are only
possible when relatively sophisticated laboratorystress-strain tests using highly undisturbed samples and/or
careful field tests are performed. This essential factor isalso elaborated in this Study.
Adopting a sophisticated automated triaxial testing systemwith the capability of measuring axial strains of ε10
and performing single tests on single specimens subjected
to varying geotechnical conditions over a very wide rangeof strains {ε10 % to ε
20 %}, the pre-failure
deformation characteristics of relatively stiff structured
clays are evaluated for purposes of determining andanalyzing parameters for design, construction andprediction. Experimental simulation is comprehensively
undertaken regarding the possible changes that may be
impacted by varying modes and levels of loading on suchgeotechnical features as multiple kinematic yielding,effects of recent stress-strain-time history, anisotropy,
structuration, destructuration, non-linearity in respect tothe time-dependency of stress-strain behavior of clayey
geomaterials and cyclic loading in both compression andextension for purposes of enhancing the sophistication of
the retrospective analysis for modeling, simulation andprediction.

Figure 3.1 Effect of Long Term Consolidation on Initialmodulus, Emax within the Elastic Limit Strain {(εa)ELS(YI)}.

Figure 3.2 Effect of Long Term Consolidation on the
Elastic Limit Strain {(εa)ELS(YI)}.
Fig. 3.3, which simulates cyclic loading→ unloading→reloading in short term and long term consolidation shows
that: 1) the initial modulus decreases exponentially withincreasing Over-Consolidation Ratio (OCR) and ispractically insensitive to the hysteritic load ↔ unload ↔
reload cycles; 2) the initial modulus is virtually the same atconstant OCR notwithstanding creep effects (secondary
consolidation time); 3) long term secondary consolidation
has practically no effect on the initial modulus. However, it
was found to significantly increase the size of the initial
yield surface {(εa)ELS(YI)} and; 4) loading ↔ unloading ↔
reloading cycles in consolidation have minimum impact on
YI when compared at a constant OCR.
Equation (15), generated from this investigation, defines
the generalized relation between the initial shear modulus,
G0 and the Over-consolidation Ratio (OCR) for relatively
stiff to hard clays.

3.2 Influence of Consolidation Stress-strain Histories
Moderate to strong vibrations and ground motions may
cause abrupt changes in the recent consolidationstress-strain-time history transforming it to a “new current
state”. It is therefore important to evaluate such changes inorder to capture a more realistic picture of the field
behavior for purposes of determining the relevant and mostappropriate design parameters. Simulation of the impact ofsuch changes is made in Figs. 3.1 ~ 3.3.
It can be noted from these figures that, although the initial
deformation resistance defined in terms of Elastic Yield
51

(15)

in the range of a >0.002%/min and, 3) through thewhere, ( G 0 )  ss is the initial modulus at the designatedcomparison of drained and undrained initial stiffness, it
can be noted that the difference becomes more significant{(Eeq)u - (Eeq)α} as the strain rate becomes exceedingly
slower, confirming the effects of time-dependent viscosity
even in the region of very small strains and the existence ofan Elastic Limiting Line (ELL) for slower rates of axial
strain; 4) the effect of strain rate is clearly significant in the
YS region ε
5 10 % , whereby the YS locus
5
expands with increasing strain rate; 5) for ε
10 %, when the strain rate a <0.03%/min, the damping

G0  {(G0 )ss / GR  OCR 0.39 }  (G0 )ss .

stress point and G R  290 MPa is the reference initialshear modulus as determined in this Study.

ratio is seen to increase as the strain rate decreases, acharacteristic which may be attributable to the creep
effects associated with smaller equivalent Young’s
modulus at slower loading rates. Essentially therefore, it
can be considered that even at very small strain rates,
viscosity cannot be ignored and that, as suggested by [11],[12], an upper bound limit may exist for Eo or Go values for
extremely large strain rates, acting as an envelope for an
ELL for the initial stress~strain relations in YI zone.

Figure 3.3 Effect of OCR and LTC on initial modulus.
3.3 Impact of Loading Rate on Small Strain
Parameters
The effect of viscosity on small strain characteristics ofvarious geomaterials has been a topic of study for anumber of researchers. Although most results have shownthat viscous effects may be minimal in the region of verysmall strains, whether or not they can be ignored has still
been debatable.
In order to investigate the effects of strain rate on the small
strain behavior of relatively stiff to very stiff undisturbed
Pleistocene clay, CUTC tests were undertaken on Osaka
Bay and OAP clays under quasi-static-monotonic loadingwith small load ↔ unload ↔ reload cycles within therange of 0  a 0.005% at different strain rates ( a ),Figure 3.4 Summary of small strain cyclic behavior for asingle amplitude ε
5 10 % {(εa)SYS(YS)} forvarying strain rates, 0.000068%/min a 0.34%/min).

part of the results of which are summarized in Fig. 3.4.
CUTC/E and CDTC/E cyclic loading through sevenέ
0.34%/min forvarying strain rates 6.8 10
single amplitudes of ε
5 10 % and ε
5 10 % were also performed.
At every stress state and small strain shear stage, the creep
rates were monitored during the secondary consolidation
stage until they attained a minimal value of a

3.4 Geo-changes Due to Varying Stress States in
Reconsolidation and Destructuration
In this section, the impact due to varying “current stressstates” and that of destructuration on some of the natural
well cemented and highly structured Pleistocene andHolocene clays found in Japan, is analyzed inconsideration of the change in the geotechnical state andbehavior in retrospect to, and as a result of, moderate tostrong vibrational and/or seismic ground motion. In sodoing, four causes of structural destruction namely;
swelling, light densification through cyclic prestraining,heavy densification through SHANSEP reconsolidation
and remolding through reconstitution, are considered. It is
important to note that in Japan, undisturbed saturated
Pleistocene clays usually exist deep in the ground, where
the confining pressure is relatively large, making themhighly sensitive to destructuration.

=0.0058%/day basically with the objective of ensuring thatno interaction occurs between the creep rates and theslowest strain rate which was designated after preliminary
investigations were carried out at a creep strain rate of a=0.098%/day ( a =6.8

10 %/min).

It can be derived from the results in this figure that: 1) the
characteristics per loading cycle are practically similar andindependent of Strain Rate (SR) even when the slowest SR,
a = 6.8 10 %/min and the fastest SR, a = 3.4
10 %/min (5,000 times faster) are compared; 2)dependency of the initial stiffness, Eeq becomes slightly
more significant as the strain rate becomes much slower,
whereas it is practically insensitive once the loading rate is(1) Stress states for varying reconsolidation regimes52In order to investigate the influence of reconsolidationstress path and regime on the “current” stress states,
specimens of OAP clay were reconsolidated to various
stress states and axial stresses equivalent to the fieldoverburden pressure (σ ) tracing different stress paths.
Fig. 3.5 shows varying stress states at which the small
strain deformation characteristics were investigated by
performing small cyclic TC probing tests. Fig. 3.6 depicts
the influence of current stress state-induced anisotropy on
the normalized Elastic Limit Strain {(εa)ELS(YI)} for bothnormally and over- consolidated specimens.
The results in this Study showed that the initial shear andYoung’s modulus as well as the YI zone are stress state
dependent notwithstanding the consolidation stress ratio,
when compared at the same stress state and magnitude.
Furthermore, the size of the initial yield surface YI exhibitsinteresting characteristics, whereby upon virgin loading atconstant p' (stress points 1→2 →3), YI expands in size.
However, once it is unloaded and offset (stress points
6→7→8) and reloaded at constant p' (stress points 8→9),
YI is compressed, contracting in size. When compared toother results from the same Study which are not reported
herein, consistent expansion of YI with Kc loading and Ko
rebounding was observed. It is considered that, thistendency, perhaps as a result of the memory of the mostrecent history, seems to persist in Ko compression (stresspoints 9→11), only to recover through Ko reboundindicating that the initial yield surface is significantly
influenced by the current stress state-induced anisotropy.

(2) Influence of destructuration due to swellingFig. 3.7(a) compares the reconsolidation and undrained
shear stress paths for intact LTC, intact STC and LTC
subjected to swell for OAP clay specimens. The three
specimens were all reconsolidated through a similar stress
path to the in-situ state of stress σ
690kPa andequivalent OCR conditions (OCR=1.86). On the other
hand, Fig. 3.7(b) shows the pre- and post- failure large
strain characteristics of the same.
It can be noted from these figures that destruction throughswelling affects the shear stress characteristics in the p'–qplane, impacts on the overall stress~strain behavior andcauses significant reduction in the maximum shear
strength notwithstanding the effects of relative ageing(secondary consolidated effected for 376Hrs.), as
discussed in the preceding sections.

Figure 3.5 Stress paths during reconsolidation and, stressstates at which small strain Cyclic Loading (CL) and Cyclic
Prestraining (CP) were performed during drained andundrained shear.

Figure 3.7 (a) Reconsolidation stress paths and, (b)Deviator stress-axial strain relations for ε4%
On the other hand, the stress - strain characteristics in the
YS zone for pre-swelling and post-swelling behavior are
shown in Figs. 3.8 (a) and (b) respectively. It can bederived that swelling contracts the YT, YS and YI sub-yieldsurfaces and causes reduction in the initial modulus anddeformation resistance. It can therefore be concluded, byinference that, swelling causes damage to the inherent
structural components such as bonding, particle matrix,
water cation, thixotropic and creep related cementation,
particle agglomeration and orientation, among otherintrinsic geo-properties. Comparison of Figs. 3.8 (a) and(b) indicates that restructuration was not achievablethrough ageing for a period of 376Hrs. of secondary
consolidation.
Nevertheless, it can be observed that there is a tendency of
structural recovery with ageing. Such recovery, as
determined from the GECPRO model, would require a

Figure 3.6 Stress induced anisotropy: Effect of varying“current stress” states on Elastic Limit Strain {(εa)ELS(YI)}.
53

magnitude of deformation resistance decreases
significantly with increasing degree of SHANSEP.
It can therefore be concluded that reconsolidation to ahigher stress level exceeding the yield stress level σ
destroyed the inherent structure of the natural, sensitiveand highly structured Holocene Ariake clay.

secondary consolidation period of at least 9,897hrs(412days) to achieve maximum stiffness and 32,859hrs(1369days) to recover the original size of the initial yieldsurface aged at constant consolidation stress state
690kPa and OCR=1.86. Whether
conditions of σ
the structural components can be fully recoverable is amatter that certainly requires further comprehensive
research.

Figure 3.9 (a) SHANSEP reconsolidation effective stresspaths, and (b) Normalized small stress-strain {(εa)SYS(YS)}
0.01% .
behavior εFigure 3.8 Effect of LTC on the small strain {(εa)SYS(YS)}
0.01% for: (a) pre-swelling and, (b)behavior εpost-swelling conditions.

(4) Destruction caused by reconstitution
An experimental regime was designed to study the
behavior of reconstituted Pleistocene clay that was, in its’
initial state, naturally well cemented and highly structured
stiff overconsolidated OAP clay. The results comparing
stress – strain behavior and the strain level dependency ofthe stiffness of the intact and reconstituted specimens areshown in Figs.3.10 and 3.11 respectively.
A number of derivations can be made from the results inthis Study including: 1) it was observed that the specimens
trace different shear stress paths right from the small strain
region throughout to failure and post-failure; 2) theyielding characteristics are quite different between the two
specimens; 3) the stress path of the reconstituted specimentranscends to a critical state at a stress ratio { μ, which is well below that
1.319, where μq/p
1.636 and, 4)located by the intact specimen μ
reconstitution caused a large drop in the inherent suction
stresses, shear strength and stiffness. These observations
imply that factors other than the decrease in void ratio such(3) Structural damage caused by over-densification
The SHANSEP reconsolidation procedure was applied on
two specimens, TS3-2 and TS3-3; whose characteristics
were compared with those of a Normally Consolidated
Intact (NCI) Specimen T-9-3 of the Ariake clay as shownin Fig. 3.9 (a).
The small stress-strain results are shown in Fig. 3.9 (b). It
can be noted that, as a result of structural destruction: 1)
the SHANSEP specimen exhibits excessive deformation;2) the peak shear stress ratio of the SHANSEP specimens
is much lower, i.e., [ q/p= 0.328] < [ q/= 0.456 < [ q/p= 0.689].
p
As a result of different structure formation and/or degree ofdestruction, the following derivations can be made fromthese figures; 1) all three specimens exhibit very different
stress-strain characteristics at all levels of strain andmultiple kinematic sub-yield surfaces; 2) the influence ofthe different stress-strain-time histories experienced
during reconsolidation can be clearly observed; 3) the54

as the breakage of the bonds between particles are the1.49, while e
1.371 .
attributes; e

4. CHARACTERIZATION OF BLACK COTTONSOIL
SUBJECTEDTO
TRAFFICTYPE
VIBRATORY LOADING
Traffic induced subsurface stress can be characterized bydynamic cyclic loading impacting numerous cycles andcontinuous rotation of stress axes that cause subsidence
and progressive distortion as well as deformation once thestresses and strains overshoot the initial yield subsurface.
While investigating the contraction of soil subjected totraffic-type stress application by applying strip loading
idealization of a moving train, [15] concluded that cyclic
stress induced by travelling strip loading generates aground subsidence that is twice as large as that observed
from cyclic triaxial compression tests i.e. (Δε)TL=2(Δε)CT.
They attributed the anomaly to stress axis rotation.
In this Study, twelve full-scale field experimental testing
sections were designed and constructed with six varying
pavement layers and structural configurations founded on
Black Cotton Soil (BCS), and subjected to different modes
and cycles of static and dynamic loading [16]. The
sub-base and base course layers, constituent of lateritic
soils, were stabilized by applying the OBRM (Optimum
Batching Ratio Method) and the OPMCS (Optimum
Mechanical and Chemical Stabilization) technologies
respectively [17]. In analyzing their effect on the BCS
subgrade and upper layers, the equipment was considered
to be essentially symmetrical about a longitudinal axis andthe wheel loads were summed up to give axle loads whichwere adopted as nominal values in the computation andanalysis of their static impact on the layers of construction.
The dynamic component however, was simulated as
moving constant wheel loads whose transfer mechanisms
occur very slowly relative to the frequencies involved in
the vertical dynamics of the equipment. In carrying out
these analyses, the mid elements of each layer are adopted
in relation to the applied stresses. Fig. 3.12 depicts
deformation characteristics simulating deterioration in
deformation resistance due to the effect of dynamic
vibratory loading on stabilized and non-stabilized BCS
subgrades.

Figure 3.10 Overall stress-strain relations for intact and1% }reconstituted specimens {(εa)T,FYS(YT,F) εFigure 3.11 Strain level dependant decay curves for intact
and reconstituted specimens, {(εa)ELS(YI)} ~ {(εa)FYS(YF)}.
A comparison of the undrained shear characteristics
between the intact and reconstituted specimens shows thatthe destructuration during reconstitution caused: 1) a
639kPa anddrastic drop in the peak strength { q334kPa, while normalization accounting forq= 0.319 andvoids ratio effects results in q/p= 0.107]; 2) conspicuous strain softening
q/p(post-failure brittle behavior) for the intact specimen
compared to the ductile behavior of the reconstituted one;3) the stress-strain behavior in the intermediate strainlevels is significantly different; 4) the intact specimen
exhibits less non-linear response; 5) the maximum
stiffness of the reconstituted specimen, E= 287MPa=
is much lower than that of the intact one, E
634MPa; 6) the intact specimen exhibits better recoverable
behavior in the region of very small strains and, 7) the
intact specimen exhibits a much higher level of
deformation resistance.
In general, these results confirm that, whilst the intact
specimens sustained most of their structural components
from the region of very small strains throughout thepre-failure zone up to the post-failure plastic yielding zone,
the reconstituted ones seem to have attained a totallydifferent “structure” having completely lost the inherent
and intrinsic structures.

Figure 3.12 Deformation resistance depicted in terms ofstrain vs. dynamic loading cycles (converted axles) tested
on specimens after 7 days cure of Western Kenya Black
Cotton Soil55

From Fig. 3.12, it can be derived that: 1) after 100,000
converted axles, the deformation of the non-stabilized
BCS specimens increases drastically (it collapsed after
336,000 cycles); 2) prolonged curing increases
deformation resistance of stabilized BCS immensely, 3)
the magnitude of plastic straining is largely reduced bystabilizing BCS, and by subjecting it to controlled dynamic
loading within the initial yield surface {(εa)ELS(YI)} whilst
applying very small amplitude pre-straining [16].
Comparison of the stress - strain characteristics of BCS
and the Japanese Kanto Loam is made in Fig. 3.13. The
specimens were initially subjected to dynamic loading andsubsequently loaded under static conditions up to failure. It
can be observed that, notwithstanding the magnitude, both
BCS and Japanese Kanto Loam exhibit a significant
increase in strength with increased initial dynamic
pre-straining at very small strain amplitudes.

6.

The author is highly indebted to Prof. Fumio Tatsuoka
formerly of the University of Tokyo under whose guidance
this research was initiated. Special appreciation isextended to the Japan International Cooperation Agency(JICA), Taisei Corporation, Japan, and Kajima
Corporation, Japan, for sponsoring most of this research aswell as the Materials Testing & Research Department of
the Ministry of Roads, Kenya and members of the Kenya
Geotechnical Society Research Team for their support.
7.
[1]

100,000
cycles
50,000cycles
10,000cycles1,000
cycles0 cycleAxial Stress (kN/m2)

800
600

0 cycle400

168,000
cycles336,000
cycles200

Japanese
Loam
Wi = 27.3%
gd = 1430
kg/m3
SD = 130
kN/m2
Kenyan Black
CottonSoil
Wi = 12.8%
gd = 1684
kg/m3
sD = 91.7

For 0 cycle only
Wi = 16.5%
g d = 1701 kg/m3
s D = 91.7 kN/m2002 Strain
4 (εa) (%)
6
Axial

8

Figure 3.13 Deformation behavior in the post-linear elastic
zone {(εa)ELS(YI)} and depiction of cyclic pre-straining
effects on the characteristics of Western Kenya Black
Cotton Soil and Japanese Kanto loam.
5.

REFERENCES

Mukabi JN, “Derived empirical relations and models of vital
geotechnical engineering parameters based on geophysical andmechanical methods of testing”, in Proc. 15th African Regional
Conf. on SMGE, Maputo, 2011, pp. 320-329.
[2] Mukabi JN, “Deformation characteristics at small strains of clays intriaxial tests”, PhD Thesis, Institute of Industrial Science,
University of Tokyo, 1995.
[3] Mukabi JN, “Innovative laboratory and in-situ methods of testing ingeotechnical engineering, to be published in the Proc. of the Int.
Conf. of the Inst. of Engineers Kenya , Nairobi, 2013.
[4] Mukabi JN, Tatsuoka F, “Effects of consolidation stress ratio andstrain rate on the peak stress ratio of Kaolin,” the 27th Annual
meeting of the JSSMFE, Kochi, 1992, pp. 655-656.
[5] Mukabi JN, Tatsuoka F, “Influence of reconsolidation stress history
and strain rate on the behaviour of kaolin over a wide range ofstrain,” Proc. 12th ARC on ISSMGE: Geotechnics for Developing
Africa, Durban, 1999, pp. 365-377.
[6] Mukabi JN, Tatsuoka F, “Effects of stress path and ageing inreconsolidation on deformation characteristics of stiff natural clays”,
Proc. 2nd I.S Pre-failure characteristics of geomaterials, Torino,
1999, vol. 1, pp. 131-140.
[7] Mukabi JN, Kotheki S, Mathematical derivative of the modified
Critical State Theory and its application in Soil Mechanics. Procs.
2nd International Conf. on Applied Physics & Mathematics, 2010
IACSIT, Kuala Lumpur, pp. 484-492.
[8] Murakami S, Hayashi S, Ochiai H, Umezaki T, “A stress parameterfor the behavior of clay under cyclic loading”, Proc. 1st Int. Symp.
on Pre-failure Deformation of Geomaterials, Sapporo, 1994, pp.
477-482.
[9] Yeoh CK, Airey D, “Undrained cyclic loading of a cemented sand”,
Proc. 1st Int. Symp. on Pre-failure Deformation of Geomaterials,
Sapporo, 1994, pp. 95-100.
[10] Mukabi JN, “Application of Consolidation and Shear Stress Ratio(CSSR) concepts in foundation design and construction” Proc. 14th ARC
on SMGE Vol. II, Yaoundé, 2007, pp.172-193.
[11] Tatsuoka F, Jardine RJ, Presti DL, Benedetto HD, Kodaka T,“Characterizing the pre-failure deformation of geomaterials”. XIV
IC on SMFE, Theme Lecture, Hamburg, 1999, pp. 2129-2164.
12 Mukabi JN, “Behavior of clays for a wide range of strain intriaxial compression”. Msc. Thesis, IIS, Uni. of Tokyo, 1991.
[13] Sica S., de Magistris F.S., Vinale F, “Seismic behaviour ofgeotechnical structures”, Annals of Geophysics, 2002, Vol. 45, N. 6,
pp. 799-815.
[14] Tatsuoka F, Shibuya S, “Deformation characteristics of soils androcks from field and laboratory test”, Keynote Lecture, 9th AsianRegional Conference on SMFE, 1992, Vol. II, pp.101-170.
[15] Towhata I, Kawasaki, Y, “Contraction of soil subjected totraffic-type stress application”, Proc. 1st Int. Symp. on Pre-failure
Deformation of Geomaterials, Sapporo, 1994, pp. 305-310.
[16] Mukabi JN, Kotheki S, Ngigi A, Gono K, Njoroge BN, Murunga
PA, Sidai V, “Characterization of black cotton soil under static anddynamic loading”, Proc. Int. Conf. on GE, 2010, Tunis, pp. 67-81.
[17] Kogi SK, Mukabi JN, Ndeda M, Wekesa S, “Analysis of enhanced
strength and deformation resistance of some tropical geomaterials
through application of in-situ based Stabilization techniques”, to be
published, Proc. 1st Int. Conf. on GEOMAT, Mie, 2011.

Comparison of Dynamic Loading Effects on Kenyan Black
Cotton Soil and Japanese Kanto Loam

1000

ACKNOWLEDGMENT

CONCLUSIONS

Overall, the following conclusions are derived from thestudy reported in this paper.
1) The conventional Critical State Theory is inadequate ineffectively modeling the behavior of well cemented
and highly structured natural clays.
2) The MCST introduced in this Study may be useful as abasis for further research, retrospective analysis,
simulation, prediction and performance based design.
3) Practically all the results in this study show that,
although the properties in the region of linear elasticity
are virtually insensitive to the effects of consolidationstress-strain-time histories, and loading rate, they arehighly susceptible in the subsequent sub-yield surfaces
in the region of small to moderate strains.
4) The magnitude of the initial modulus and size of
sub-yield surfaces are “current” stress state dependent.
5) Remolding (reconstitution) completely destroys theintrinsic structure of structured clays. Theories andconcepts based on tests performed on reconstituted
clay specimens cannot therefore model the behavior ofnatural clays.
56

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

The Maintenance of Ground Anchors in Nippon Expressway
Yuu FUJIWARA,Nippon Expressway Research Institute Company Limited
Toshinori SAKAI, Mie University


ABSTRACT: On expressway in Japan since around 1970, theground anchor (the anchor) has been adopted. Since then,construction has been increasing in the mountainous road, which
increases the slope of the cut earth stabilization measures for
Industry and Labour landslide measures. This has increased theamount of construction of anchor every year. And that number is
now more than 120,000 books. Anchor performs maintenance on aregular basis, you need to make sure it is functioning properly. In
particular, make sure to keep tension force is introduced to theanchor is important. How to determine the tension force is twofold.
One is how to measure and install a load meter anchor. Second, there
is a Lift-off test tensile load is applied directly to the anchor. During
shared lift often tested. But, however, Lift-off test, no uniform
standards have been done differently at each site. There is also a
problem that takes time to conduct the test. Therefore, efforts toimprove the efficiency of test methods and testing concrete.

Fig.1 Groundanchor
receivving pressure plateanchorhead

Keywords: groundanchor,Lift-off test,expressway,soil
1. INTRODUCTION
Ground anchors (hereinafter referred to as anchors) are acountermeasure of stabilizing against the landslide and cut
slope(Fig.1).
Anchors are composed of anchorhead and piston plate
exposed on the surface, and tension member and anchorbody
built in the ground(Fig.2)．Construction of the anchor, theanchor reclamation to ground the body is done in a way thatfixing the head was anchored in a state of tension force to thetension, this tension is exerted on the pull. In the mid-1960s
the highway was first adopted by ground anchors. Thereafter,
the annual increase with an increase in the amount ofconstruction of road construction in mountainous regions,which now number more than 120,000 books(Fig.3).
The construction of the initial anchor, the water and air
infiltration due to insufficient corrosion features on theperimeter of the body pulling the anchor and the anchor heador anchor, and a structure susceptible to progressive
degradation of the material tensile other. In late 1975, now
seen as symptoms of lifting the anchor and jump out, JH Old
for This (NEXCO present), the regular aspects ofconstruction law has been anchored since around 1985 havebeen working on making the inspection and maintenance.
Society in 1988, the old soil (JGS current) standards havebeen revised, the anchor started to be adopted and improved
corrosion protection capabilities. Has been applied to the
local design guidelines based on the criteria established in
1992 in this former JH. Thereafter, the implementation of
emergency measures and the anchor steel bar, in order toimprove assessment accuracy and efficient execution ofsurvey work and health inspections anchor, hammering due toperiodic inspection and detailed visual inspection due to daily
inspections and and to confirm that status. In this paper, we

Fig.2 Groundanchor’s structure（×1,000km）
8

アンカー施工本数の累計
construction number
total of exressway use
高速道路供用延長の累計
extension

100
80

6
4

60
40

2

20

0

0
Ｓ44
1969

Ｓ52
1977

Ｓ60
1985

Ｈ5
1993

total of expressway use extension

total of anchor construction number（×1,000books）
120
total of anchor

Ｈ13
2001

Fig.3 Anchor’construction situation in expressway

describe the contents of the initiatives on the anchor in themaintenance of highways.
2. MAINTEANCE OF THE ANCHOR
Inspection of the anchor in the highway, depending on its
purpose inspection items, inspection methods, inspection
frequencies have been established(Table 1).
Daily inspection, for example, corrosion of materials Status

57

Table 1 Anchor's check typecheck type

Initial check

Daily
checkupRegularservice

Detailed
check

content
Check that is evaluated and judged bymeasuring adjacent watching, slapping
sound, and load to see how matters stand in
the early after structure is completed, and tounderstand progress condition of thefollowing damage adequately before it
begins to use it.
It checks by adjacent watching or.hope.
watching getting off if necessary to finddamage of the anchor placing region of the
management district whole etc. at the earlystage based on watching on the car. The
fourth/two weeks. It does at the frequencyof the seventh/two weeks.
Check done to understand situation ofchiefly individual anchor of anchor
installation slope based on adjacentwatching on foot. It does at the frequency
once/year.
Check done to execute adjacent watching,slapping sound, and other concrete checks〈 measurement and load measurement 〉
of anchor installation slope on foot, and tounderstand situation of chiefly individual
anchor. Frequency.do.

Temporarycheck

Check done if necessary in supplementation
and abnormal weather etc. of daily checkup.

Soundnessinvestigation

Investigation that observes, checks headestablished body in detail in the one done
when damage of strangeness etc. areconfirmed in anchor installation slope, andfull investigation is needed, and measuresload in addition.

Fig.4 Anchor load managementload(P)remaining hitching
power(Pe)anchorheaddisplacement(δ)

Fig.5 Lift-off test

because of the slope by excavating to install a load meter forsome of the anchor, we will proceed to construction, making
sure to close the residual tensile strength. In-service for theanchor, the load management is done on a regular basis in a
way to measure the residual tension load cell was installed atconstruction. And the total comes to be degraded over time
due to load, however, usually is very difficult load meter
replacement, etc., will anchor the remaining tension in the
Lift-off test conducted to measure.
2.2 Lift-off test
Lift-off test, a hydraulic jack installed on the head portion of
the extra length of the anchor tension member is performedby loading a tensile load. Figure 5 shows the status of theLift-off test in fixing wedge type anchors. Fixing wedge
Lift-off test of the type anchor, the anchor head portion of theextra length, pooling the head provisional established tosecure the tension bar and the extra length portion is
performed by loading a tensile load to the tension bar. In theprocess of loading, the residual tension when the anchor
isconfirmed, the unloading of the load carried into the wild.
Figure 6 shows the state of the anchor head in the process ofloading unloading Lift-off test. In addition, a general load isdrawn at the time - the shape of the curve in Figure 7 shows
the displacement. In the process of loading the Lift-off
testThe stage, before starting off the head anchor plate,tension anchor fixing and visual equipment, lifting andpeeling to confirm the availability of protective concrete byhammering. If you notice unusual changes in the shape of thevarious inspection results, a survey to examine the health of
the anchor. The study is detailed observation of the bodyfixing Verification head to remove the protection work of theanchor head, measure the residual tensile force necessary toevaluate the soundness of the anchor.
2.1 Residual tensile strength of anchor Management
Upon maintaining the anchor, the tension continued toinvestigate the residual, that is important to continue to takeappropriate action. The anchor introduced the ability to settle
when the initial tension, tensions may weakened slightly cut
into the material loss due to set tension on the anchor head.
Each method considered in this loss of the anchor set, theintroduction of the initial tension force so that force is slightly
lower than allowable tension force when fixing anchors.
When fixing the anchor, the residual tensile strength of theanchor tends to decrease due to soil creep and relaxation oftension wood, and will gradually converge to a constantvalue.
Anchor for the highway, Figure 4 shows the flow of thegeneral management of the residual tensile strength. In theconstruction phase, the ground Instability is a concern
58

Pe

P=0

P

δ

Pe

processes except load

process of applying load

（Ⅰ．hitching load doesn't act）

P<Pe

（Ⅱ．state before off lift）
P

δ

Pe

P=Pe

（Ⅲ．hitching load≒remaining hitching power）

Fig.9 Comparison with examination in the past

δ

P

Pe

Table 2 Result of lift off examination using SAAM jack

P>Pe

（Ⅳ．state after lift‐off）（P ： load of hydraulic lifter，Pe ：remaining hitching power）
Fig.6 When the tension load is put and when it excludes itload(P)D

process ofapplying load
（flexion point）

E

processes
except load

C

B

A
O

F

displacement(δ)

residual displacement

Fig.7 by Lift-off test load - displacement curveloading (Fig. 5; Ⅲ, -6 Figure; point A). This phenomenon is
generally called "lift" is called a load - a measurement of theload-displacement curve at the time the lift-off "the lift" and it
is. And further increase the tensile load was lifted off and thenbegin to gradually change in the slope of the linear gradient
between OA (-5 figure; Ⅳ, -6 figure; A → B). Once thatchange is constant, the tensile load transmitted to the anchor
free length portion, the load - displacement curve shows agradual linear slope depends on the tensile modulus of thematerial (Fig. 5; Ⅳ, -6 Figure; B → D). Residual tensile
strength of the anchor, JGS criterion 5), according to the headstarted away from the anchor plate (0.1 ~ 1.0mm) has beenmeasured at the load. After the loading of the tensile force
necessary to evaluate the residual tensile strength of anchor

Fig.8 SAAM jackanchor head displacement is small, the load - displacement
curve shows a linear gradient of steep slope (Fig. 5; Ⅱ, -6
figure; O → A). Increase the tensile load of hydraulic jacks
even begin to anchor the head away from the plate
immediately after the load was equal to the residual tension
59

loading to unloading the load (Figure 6; D → E → F). In
addition, after completing one cycle of loading unloading, theresidual displacement is the difference between before andafter loading displacement (Figure 6; OF) may occur.
Residual tensile strength, if reduced to less than 90% of time
fixing force or tension increased over 120% of the anchor
design strength, may not be anchors healthy functioning.
3. Efforts to improve the maintenance of anchor
Lift-off test, set a foothold in the cut earth surface For the
implementation method Only in some cases you need to do atest using a large crane. Because of the limited test volume
and number of test cases will require significant cost and time.
SAAM has been developed in the laboratory jack Mie Sakai(Fig. 8), may be to generate a large tensile strength
lightweight and compact construction. Therefore, since iteliminates the need for such a temporary scaffolding was
required that the conventional test, it is possible to quickly
investigate a number of anchors (Figure 9). The condition canbe performed more than 20 per day, it is possible todramatically extend the enforcement decree of the day.
SAAM test is carried out by the jack lift anchor and head forthe slope of the study six of No.1 ~ No.6, examined the
distribution of residual tension strength of the anchor (Table
2).
About 137 tests were carried out this anchor lift / day on
average for six of the slope gradient is 1:0.5 to 1:0.8, wereable to present an average of 16 in 1:1.0 to 1:1.2. For the
implementation of good work at the point where the slope
gradient is testing many loose number of tests per day, could
be carried out up to 24. Construction is considered asignificant capacity Given that this was about 2-3 per day of
conventional test methods.
Currently, NEXCO, they can be applied in the field trials
utilizing the SAAM lift jack, which repeated studies.
4.CONCLUSION
Anchor conducted after the construction and maintenance, it
is important that you make sure it is functioning properly.
However, the situation has not been adequate maintenance ofeffort it takes to be a huge amount of maintenance andconstruction.
Future, by further analysis of the data accumulated in
NEXCO, efforts shall be made to establish the techniquemore efficient and effective maintenance.

5. REFERENCES[1]

JGS, ground anchor design and construction standards, the samecommentary, 2000

60

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Stress Distribution in Khon Kaen Loess under Spread Footing
Pairoj Yodsa-nga and Watcharin Gasaluck Department of Civil Engineering, Faculty of Engineering, Khon Kaen
University, Thailand.
program is necessary. Difficulty and price of the program
obstruct engineers to use it.
The research presented in this paper was done to compare thestress distribution in Khon Kaen loess due to the load acted on
0.3x0.3 m steel plate with the outputs obtained from classicalmathematical models and PLAXIS 3-D, the computer
program based on finite element method.



ABSTRACT: Loess is the non-plastic fine grain soils distributed in
Northeast Thailand. The stress distribution in loess due to surcharge
load was studied. Plate bearing test was carried out at 3-m depth inloess. Five small total pressure cells were installed for getting thestress increment in soils during the test. The test results werecompared to the results calculated by 3 mathematical models.
Boussinesq’s method and finite element method gave the outputs
those were not agreed with the test because linear stress-strain
behavior was utilized. Walter’s method provided the better output.
This method does not accept the solid mechanic assumption but
non-linear stress-strain behavior was employed. Both the test andWalter’s method show the change of influence factor along with thechange of load. Finite element method may give the reasonable
output if the appropriate model condition is specified. The contactpressure pattern should be considered as well. All mathematicalmodels used the uniform pressure pattern that some researchers did
not agree with. The actual pattern should be studied and used in themathematical models. The better result might be obtained.

2. SITE CHARACTERISTIC
Northeast Thailand comprises an area of about 168,800
square kilometers or about one-third of the total area of thecountry. Some soils in Northeast Thailand are silty sand,clayey sand, sandy silt, clayey silt, sandy clay and silty clay of
the Quaternary age, with an overlying bedrock of sandstoneand shale of Mesozoic age [4].
Khon Kaen Province is located in Northeast Thailand. The
loess deposits exist above the water table in abundance inKhon Kaen. The thickness normally ranges from a few tomore than six meters. Fig. 1 is the distribution of loess inNortheast Thailand. The loess characteristic is non-plastic red
sandy silt or silty sand (ML or SM). Most of the soil particles
are 0.005 to 0.042 millimeter in size. Soil grains have asmooth and sub-rounded surface. The microstructure isunconsolidated and loose to medium dense. The pore sizes
are usually 200-500 micrometers although some can be aslarge as 1 millimeter. Small clay content is presented in theform of clay bridge bonds [5]. Undisturbed Khon Kaenloess’s photograph taken by Scanning Electron Microscope isshown in Fig.2. Under the 90 percent standard Proctor test
condition, it is classified as slightly to moderately dispersive,
ND3 [6].

Keywords: Loess, Stress Distribution, Boussinesq’s method,
Walter’s method, finite element method
1. INTRODUCTION
Stress distribution in soils is very important in foundation
works. Not only the design but also the analysis of foundation
settlement needs this information, especially the stress
distributions of spread footing group that the overlap ofpressure bulb may cause the failure.
The classical Boussinesq’s method invented in 1883 isusually used for estimating stress distribution in soils massdue to surcharge load[1]. It is the mathematical model that thesoil is assumed to be semi-infinite solid and its strengthparameters are not taken in to account. In other words themodel assumption is doubted that soil type has no effect onstress distribution behavior. Contact pressure between rigid
footing and soils was assumed to be constant. Hansbo [2]made the conclusion of the contact pressure in clay, silt andsand studied by some researchers, only the rigid footing on
ground surface gave the contact pressure that agreed with
Boussinesq’s method.
Walter [3] developed the mathematical model for estimatingstress distribution in soil mass. Compression and shear
modulus of elasticity were the important parameters in themodel. Furthermore, the relationship between compression
and shear modulus of elasticity was different from solidmechanic theory. The output reasonably shows the effect ofsoil characteristics on stress distribution behavior.
Finite element method is becoming popular along with thefast development of computer. It is the powerful device that
could be used for any complicated soil model. The problem offinite element method is its difficulty, and then commercial

Fig. 1 Distribution of loess in Northeast Thailand.
61

distributed load acted on the ground surface. Fig. 4 shows themeaning of the parameters used in those equations. I is called
influence factor.
σ z =qI σ(1) 2mn  m 2 +n 2 +11/2
m 2 +n 2 +2 
2 22
1  2 2+n 2 +1 Iσ =  m +n +m n +1 m 1/2
4π
2mn  m 2 +n 2 +1
-1
 + tan

2
2
2 2


m +n -m n +1
B
m=
z
L
n=
z

Fig. 2 Undisturbed loess.

(2)

(3)
(4)

3. METHODOLOGY
3.1 Field and Laboratory Tests
Plate bearing test was done at the 3 m depth in Khon Kaenloess. The plate size was 0.3x0.3 m, the thickness was 2.5 cm.
Load was applied to the plate by the hydraulic jack acting
against dead load. Five 3-cm holes were bored, earth pressure
cells were placed at the bottom. The holes were filled withloess in many layers; each layer was about 10-cm thick. The
density of the filled loess was about the natural density ofloess around the holes, it was controlled by the loess weightof each layer. Fig. 3 shows the position of installation. Soilsample was collected and tested for getting its parameters inlaboratory. Triaxial tests were done on the undisturbed
samples with natural water content to get the strengthparameters.

L

Uniform
Distributed
Load q

X
B
z

Y
P
z
Fig. 4 The parameters used in Boussinesq’s equation.
Walter’s method is quite complicated. Equation (5) is thevertical stress at any point P shown in Fig. 5. The values of c2and  can be found by (6) and (7). E, G and  are compression
modulus of elasticity, shear modulus of elasticity andPoisson’s ratio respectively. In (5) z1 is the solid angle
subtended at the point (x, y, z/c) by the loaded area. The solid
angle is defined as the area on a sphere of unit radiusintersected by the elements extending from the point (x, y,z/c) to the bearing area. Also, z2 is the solid angle subtended
at the point (x, y, cz) by the loaded area. There are double
values for c2; the upper and lower signs are used for c2 and1/c2 respectively.
σz =

c2 =

Fig. 3 The installation of earth pressure cells.

λ=3.2 Mathematical Models
According to Boussinesq’s method, (1) to (4) provides thevertical stress (z) at any point P in the soils under a uniform
62



q



2π c 2 -1

1-ν
ν 2E

 c θ -θ 
2

z1

z2

E
E2
E E
±
-4ν 1+ν   + 
G
G2
λ G

E2
E
E E
 m
-4ν 1+ν   + 
λ G
G2
λ G

Eν
1+ν 1-2ν 

(5)

(6)

(7)

that the pressure distributes broader in soils.
Uniform
Distributed
Load q

Table. 1 The influence factor (I) obtained from thecalculations and test.

Position

x

Y

0.5B
B
1.5B

y
z/c

G, kPa
20000
0.200
0.141
0.157

15000
0.136
0.098
0.099

Test’s results
Pressure onplate, kPa

0.400
0.240
0.179

0.500
0.290
0.190

654
0.271
0.181
0.129

2725
0.150
0.150
0.100

X
The double consolidation tests of Khon Kaen loess showed
the large decrease of void ratio due to the high pressure [9&10]. The undisturbed loess in Fig. 1 had the void ratio of about
0.8. At the pressure of about 2000 kPa, the ratio became about
0.4 as shown in Fig. 6. It could be concluded that Walter’s
method provided the reasonable result which was affected bysoils strain level.
PLAXIS 3-D is usually used under the condition of
Mohr-Coulomb Model which solid mechanic theory is
employed. This model gave the result that was quite different
from the test’s result. The example of its output is shown in
Fig. 7. Hardening Soil Model should be more suitable but itneeds complicated parameters. Boussinesq’s method also
gave the constant influence factor. The comparison between
the calculated influence factors and the tested ones are shownin Table. 2.

zZ

Walter

PLAXIS
3-D

Y

Boussinesq

X

Depth

Z

P
(x, y, z)

cz

Fig. 5 The parameters used in Walter’s equation.
Note that the solid equation, E = 2G(1 + ) is not employed inthis method. The assumption is that the shear modulus of soil
is smaller than the value obtained from the solid equation.
This assumption has been supported by many researchers
such as [7] and [8].
The soils parameters used in the mathematical models werefrom laboratory tests. Compression modulus of elasticity (E)was 52580 kPa. Poisson’s ratio () was 0.25. In case of
Walter’s equation, two values of shear modulus of elasticity(G) were used. The first value was 20000 kPa which was nearto the 21032 kPa obtained from solid equation. The second
one was the 15000 kPa which was assumed due to the
behavior of soils at medium strain level [7]. Modulus
parameters were not used for Boussinesq’s equation. The
PLAXIS 3-D used solid mechanic theory in the analysis.

4. RESULTS AND DISCUSSIONS
Table. 1 shows the comparison of the influence factor (I)obtained from the 3 methods of calculation and plate bearing
test. The test results show that the stress distribution in soils
was not constant for all plate pressure. The increase of platepressure caused the decrease of influence factor at the point
under consideration. In other words, the higher pressure madethe high strain of soils, then, shear modulus of elasticity
reduced. The smaller shear modulus of elasticity made thesmaller influence factor at the point under consideration.
Walter’s method gave the result that agreed with the test’s
result, the reduction of shear modulus of elasticity caused the
reduction of influence factor. This could be considered thatafter the pressure acted on the soils, the void ratio wasdecreased and then the soils became harder. Therefore, thesettlement induced the better distribution behavior. It means

Fig. 6 The loess under about 2000 kPa pressure.
The pattern of contact pressure between plate and soils would
account for the difference between the factors obtained from
calculations and test. This study had the assumption that thecontact pressure was uniform but some researchers reported
that the contact pressure varied in pattern depending uponsoils type as shown in Fig.8. The Khon Kaen loess having
characteristic between SM and ML should have the pattern as
between (a) and (b) of Fig. 8.
63First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

The Effect of Sheet Pile Length on the Capacity of Improved Pile Foundation on
Sand under Vertical and Moment Loading
Tuan Van Nguyen, Pongsakorn Punrattanasin
Department of Civil Engineering, Faculty of Engineering, Khon Kaen University,
Email: ngtuanhue@gmail.com, ppunrattanasin@gmail.com,

ABSTRACT: This paper is to investigate how the sheet pilelength affects the capacity of improved pile foundation. The
improved pile foundation, in this study, can be defined as a pilefoundation skirted by sheet piles around the periphery of pile cap.
A series of 1g physical modeling tests were conducted on sandunder vertical and moment loading by the way that the verticalload was applied at different positions from the pile cap
centerline. Besides the center of pile cap, four other points withincreasing eccentricities were considered. Two types of model
pile foundation were used in the experiments and their loads andsettlements were measured by using a load cell and linearlyvariable differential transducers (LVDTs). The effect of sheet
pile length on the capacity of existing pile foundation can beevaluated by the results of load-settlement relationship. The
results show that the sheet piles significantly enhances thefoundation capacity under vertical and moment loading. In
addition, the results reveal that the capacities go up with theincreasing sheet pile lengths.

Key words: sheet pile length, pile foundation, capacity,sand
1. INTRODUCTION
Enhancing the foundation bearing capacity has beendrawing a lot of researchers’ interest because it is one of
the key issues in foundation structure design. Due to thehigh application to various soil conditions, various types
and sizes of superstructures, pile foundation is usuallypriority choice to satisfy both safety and economy in caseof weak soil and great bearing loading requirement. In
order to improve the pile foundation capacity, Nguyen andPunrattanasin [1] adopted the design concept of sheet pilefoundation which was defined as a shallow foundation
with sheet piles skirted by sheet piles around the peripheryof the footing [2],[3]. Installing sheet piles around the pilefoundation which is called improved pile foundation is oneof the ways to significantly improve the pile foundation
capacities in terms of vertical, horizontal and moment [1],[4], [5]. In this paper, a series of 1g physical modelingtests was continuously conducted on sand under combined
vertical and moment loadings with variety of sheet pilelength to study how the length of sheet pile affects thefoundation bearing capacities. All tests were completedwith the same initial condition so that the experimental
results can then be properly compared and discussed.
From the results of this study, an appropriate sheet pilelength was also revealed.

In order to properly compare and discuses theexperimental results, all the tests were done with the same
initial condition.
2.1 Model Foundation
The model foundations include four identical steel piles(200mm in length and 14mm in diameter), a steel pile cap(70mm x 70mm x 30mm) and four 1mm-thick steel sheet
piles with variety of lengths. The distance from the centerto center of piles in the group is 3d or 42mm (d is the pile
diameter). The four piles are screwed into the pile cap
whose top surface has not only a central groove for purevertical load but also many grooves in order that the load
can be applied with different eccentricities. The pile capwith many grooves is shown in Figure 1. The sheet pilesof which length is measured from the bottom of pile cap
are also connected to the pile cap by screws. A successionof sheet pile lengths (Ls) of 10mm, 20mm, 30mm and
40mm or Ls/ L ratios of 0.05, 0.1, 0.15 and 0.2 were usedin this study (L: length of pile). The variety of sheet piles
is shown in Figure 2. In addition, both sides of sheet pilesurface are not glued with sand or any material, so theroughness of the sheet pile is assumed to be smooth.

Figure 1. Pile cap with grooves

2. EXPERIMENTAL WORKS
This paper reports the results of a series of 1g loading testsof pile foundation model with variety of sheet pile lengthon sand under the combined vertical-horizontal loading.
The purpose of these tests is to investigate sheet pilelength’s effect on the moment capacity of pile foundations.

Figure 2. Variety of sheet pile lengths

65

vertical settlement of the foundation. The arms are rigid
enough to ignore their own bending. Figure 4 respectively
shows the 3D schematic drawing and general view of the
combined loading system.

Beside the models of pile foundation with sheet piles at
the four sides of pile cap (PFFS), the pile foundation
model was used in this research with the aim ofcomparison. Both of them (shown in Figure 3) under
combined vertical and moment loads were embedded in
the same sand and initial condition.

(a) 3D cross section schematic drawing
Figure 3. Two types of model foundations: PF and PFFS
2.2 Loading System
The test program in this study was performed at
Department of Civil Engineering, Khon Kean University,
Thailand, by using a new combined loading apparatus
designed and constructed by Nguyen and Punrattanasin [6].
The outstanding point of this apparatus is that it can behighly applicable for any combination of vertical,
horizontal and moment loadings and displacement paths
can be generated and applied to model foundations with
precise force, position and direction. There are four mainparts in the loading apparatus: loading system, measuringsystem, control system and data acquisition system.
Besides, the loading apparatus includes a soil container
which is linearly and smoothly movable out of the loading
system by means of its four wheels moving on two rails inorder to easily prepare the soil before testing.
Two air cylinders were respectively used for apply verticaland horizontal loads on the foundation. They were
mounted on a gantry system which is clearly shown in
Figure 4. The gantry system includes a fixed part andunfixed part moving along four rails on four I section
beams by means of its eight wheels. The horizontal andvertical air cylinders were mounted respectively on the
fixed and unfixed parts. Hence, in terms of eccentricloading cases, the vertical loading system was horizontally
moved to a position in order that it can apply the desired
eccentric point on the footing. In other words, momentloading could be generated by this way. The loads and thedisplacements of the model of foundation were measured
via load cells and linearly variable differential transducers(LVDTs). Portable data logger was used to record the datafrom load cells and LVDTs through a computer. Due tothe small size of the pile cap, two arms attached to thesurface of the pile cap via screws were used to measure the66(b) General view
Figure 4. The combined loading apparatus
2.3 Soil Sample Preparation
Dry sand, the Puttaisong sand which is easily found in the
Northeast of Thailand, used in this research is uniformly
graded sand, sorted by particle sizes. The sand was poured
from a hopper into the soil container, a steel box with
inner dimensions of 80cmx40cmx80cm in length, widthand depth, respectively. Its size is large enough with noapparent influences of boundary effects [7]. In order tocreate the same initial condition for all the tests, the height
between the bottom of the hopper and deposited sandsurface was chosen to be 50 mm and periodically adjusted
to the rise of the sand surface to maintain the constantheight until the sand was fully poured into the container.
In other words, the uniform sand, of which relative density
is constant, could be achieved through this way. In
addition, pouring the sand until the container became fullmeans that there was no apparent influence from effects
caused by the bottom of the container [7]. Table 1
summarizes the properties of the sand. After pouring sand,the loading system applied the vertical load at the center ofthe pile cap to drive the pile foundation into the sand until
touchdown. In case of pure vertical loading, the load kept
applying until failure. In cases of eccentric loadings, thevertical loading system was horizontally moved to a
position in order that it can apply the desired eccentric
point on the footing.
Table 1. Physical properties of experimental sand
Type

Sand

Unit weight (kN/m3)
D60 (mm)
D30 (mm)
D10 (mm)
Uniformity coefficient, Cu
Coefficient of graduation, Cc

14.39
0.395
0.360
0.240
1.646
1.367

(a) Ls/L= 0cm

2.4 Detail of The Tests
A series of 25 loading tests were conducted and reported
in this paper. Two kinds of foundations were used in thisresearch. They are pile foundation (PF) and pilefoundation with four sheet pile (PFFS). In terms of PFFS,
a succession of sheet pile lengths (Ls) of 1cm, 2cm, 3cm
and 4cm or Ls/ L ratios of 0.05, 0.1, 0.15 and 0.2 wereused in turn (L: length of pile). Besides the centre of pilecap, four other points with E= e/B = 0.1, 0.167, 0.25 and0.333 were made for eccentric vertical loading cases (e:eccentricity, B: width of pile cap). Each foundation tookturns to be applied by a centric loading and four eccentricloadings.

(b) Ls/L= 4cm

3. EXPERIMENTAL RESULTS
A foundation under an eccentric vertical load V may beconsidered as subjected to a vertical load V and a moment
M, which is equal to the load V times its distance from thecentre. In other words, to generate the moment, the model
foundations are applied a vertical force at different
distances from the pile cap centerline. It should be notedthat the horizontal displacements of model foundations
were neglected in all experiments until the failure
conditions were approached. The results of the series aregraphically shown in Figures 5 at Ls/L ratio of 0, 0.05, 0.1,
0.15 and 0.2, respectively. In this paper, the load at asettlement of 0.1B (7mm) was chosen as the ultimate
bearing capacity of the foundation [8]. All the values ofbearing capacities were summarized in Table 2(c) Ls/L= 5cm(d) Ls/L= 6cm

67

(e) Ls/L= 7cm
Figure 5. Vertical load- settlement relationship of model
foundations with different Ls/L ratio

Figure 7. Bearing capacity of various type of pilefoundation model under eccentric loads

Table 2. The bearing capacity of all kinds of foundations
Improved Pile Foundation
with Ls/L =
0
0.05
0.10
0.15
0.20

E=0
286
312
340
365
415

Bearing Capacity (N)E=0.1
E=0.167
E=0.250
252
236
226
280
260
245
312
298
279
337
321
310
395
379
369

E=0.333
193
203
232
260
332

Table 3. The moment capacities’ value and greater percentage compared with PF of all kinds of foundations
Foundations with Ls/L =Moment Capacity
0
0.05
0.10
0.15
0.20
Value (Ncm)
527
553
681
807
1281
Greater percentage compared with PF (%)
0
4.9
29.3
53.2
143.3
Table 2 obviously presents that the bearing capacities ofthe foundations rise moderately with increasing length of
sheet piles regardless of eccentricities. This shows moreclearly in Figure 6. The increasing capacities could be
explained by contribution of (1) the skin friction along thesides of the sheet piles, (2) the end bearing of the sheet
piles and (3) the confinement effect of soil inside sheet
piles and pile cap.

As expected, no matter how long the sheet piles are,certain decreases of the capacities with increasing
eccentricity are expressed in Figure 7, where the bearing
capacity V is plotted against the ratio of the eccentricity(e) over the width of pile cap (B).
Figure 7 also reveals that the capacities enhance graduallywhen the sheet pile lengths raise and they quickly go up at
the ratio Ls/L of 0.2.
In order to determine moment capacities, the effect of the
moment acting on the foundations is plotted in Figure 8 atdifferent Ls/L ratios in terms of a normalized diagram on
M-V plan with respect to Vmax (centric bearing capacityof pile foundation) [2].

Figure 6. Eccentric vertical capacity of foundations withdifferent sheet pile lengths

68

[2] Punrattanasin, P., Kusakabe, O., Murat, O., Koda, M.,
Nishioka, H. 2002. Sheet Pile Foundation on Sand under
Combined Loading – A Literature Review and Preliminary
Investigation, Technical Report of Tokyo Institution ofTechnology (65): 57-85.
[3] Nishioka, H., Koda, M., Hirao, J., Higuchi, S. 2008.
Development of sheet pile foundation that combines
footing with sheet pile. QR of RTRI 49(2): 73-78.
[4] Takemura, J., Izawa, J., Hiroaki. and Yamana.,“Centrifuge Model Tests of Pile-Sheet Pile Combined
Foundation Subjected to Lateral and Moment Loading”,
Proceedings, 6th Regional Symposium on Infrastructure
Development, Bangkok, Thailand, 2009.
[5] Nguyen, T.V. and Punrattanasin, P., “The Capacity of
Improved Pile Foundation on Sand under Vertical andHorizontal Loading”, International Conference Advancesin Geotechnical Engineering - ICAGE, Perth, Australia,
2011 (full paper accepted).
[6] Nguyen, T.V. and Punrattanasin, P., “Development of
Combined Loading Apparatus for Physical Modelling”,
Research and Development Journal of the Engineering
Institute of Thailand, 2011, (accepted and will be
published).
[7] Al-Mhaidib, A.I. 2006. Experimental Investigation of
the Behavior of Pile Groups in Sand under Different
Loading Rates. Geotechnical and Geological Engineering
24:889-902.
[8] Punrattanasin, P., Gasaluck, W., Muktabhant, C.,
Angsuwotai, P. and Patjanasuntorn, A., “The Effect ofSheet Pile Length on the Capacity of Sheet Pile
Foundation”, Proceedings, 17th International Conference
on Soil Mechanics and Geotechnical Engineering,Alexandria, Egypt, 2009, pp 598-601.

Figure 8. Normalized load-moment relationships
Figure 8 indicates that the maximum moment offoundations (Mmax) can be approximated by Mmax/BVmax
ratio. It should be noted that Vmax is maximum centricvertical of the pile foundation (PF). From Figure 8, theMmax/BVmax ratio are 0.263, 0.276, 0.340, 0.403 and 0.64
in cases of Ls/L equals 0, 0.05, 0.1, 0.15 and 0.2,
respectively. In terms of the foundation moment capacity,
the values and the greater percentage compared with PF
are shown in Table 3 which illustrates that the longer sheet
piles are, the greater moment capacity gets. It is interesting
to note that the moment capacity also soar up at the ratio
Ls/L equals 0.2.

4. CONCLUSION
The series of twenty five physical modeling tests wereconducted on sand under a series of eccentric loadings toevaluate the influence of sheet pile length on the pilefoundation capacities. The moment capacity offoundations could be determined from a series of eccentric
capacities. The results clearly demonstrate that attachingfour sheet piles around the pile foundations provides
significant improvements in capacities both vertical andmoment. The capacities dramatically increase when theratio Ls/L reaches 0.2. In deed, the greater percentages ofcapacities compared with pile foundation are 45.1% and143.3% in terms of vertical and moment respectively.
5. ACKNOWLEDGMENT
This research was financially supported by the researchfund from Faculty of Engineering and Graduate School,
Khon Kaen University and Khon Kaen University
Scholarship for Human Resource Development to
Neighboring Countries.
6. REFERENCES[1] Nguyen, T.V. and Punrattanasin, P., “The Capacity of
Improved Pile Foundation on Sand under Vertical andMoment Loading”, Proceedings, 3rd Technology andInnovation for Sustainable Development International
Conference, Nong Khai, Thailand, 2010, pp 126-130.

69

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Investigation of a Reliable Reinforcing Method for Embankment Ground
H. M. Shahin, T. Nakai, M. Kikumoto, R. Sakai, Y. Yoshida, I. Saito, N. Nagao & K. Toda

subloading tij model [5]. This model can describe the typicalstress, deformation and strength characteristics of soils, suchas the influence of the intermediate principal stress, theinfluence of stress-path dependency of the plastic flow andthe influence of the density and/or the confining pressure.

ABSTRACT: In this research, effectiveness of reinforcing inembankment ground has been investigated and reinforcing
mechanism has been clarified by laboratory model tests and thecorresponding numerical simulations. Three reinforcing methods are
employed, namely, sheet pile alone, sheet pile with tie-rod, and acomposite method of sheet pile with nailing. The numerical analyses
are carried out with a finite element program called FEMtij-2D,using the elastoplastic subloading tij model. In the case of thereinforcing with the sheet pile alone, no reasonable increase of the
bearing capacity of the reinforced ground is observed. In contrast, inthe case of the composite reinforcing method of sheet pile withnailing, a significant increase of the bearing capacity is seen as thelateral displacement of upper part of the sheet pile is impeded due tothe inclusion of the nailing. The finite element program FEMtij-2D
properly predicts the results of the model tests on the reinforced
embankment.

2. OUTLINE OF MODEL TESTS AND ANALYSES
2.1 Model Tests
Fig. 1 shows an apparatus for laboratory model test. The
model test is conducted with the aspect ratio of 1:100 between
the model tests and prototype scale. The width of the groundis 100cm having the height of 50cm. A stack of aluminumrods, in which two kinds of round rods having diameters of
1.6mm and 3.0mm are mixed in the weight ratio of 3:2, wasused as the model ground. The unit weight of the mass ofaluminum rods is 20.8kN/m3 at model stress level. The widthof the loading plate is 12cm with the thickness of 3cm whichcorresponds to the width of the embankment in real field. The
sheet piles are emulated with aluminum boards. The thicknessof the aluminum board is obtained considering the similarity
ratio between the model tests and prototype scales. Here, twotypes of rigidities – flexible and stiff with the thickness of
0.3mm and 0.5mm, respectively, are employed. The distance

Keywords: Embankment, reinforced ground, bearing
capacity, finite element analysis
1. INTRODUCTION
Many railways and highway embankments are built on soft
soils because of the limitation of firm ground and therequirements for a straighter alignment. In these cases, it isnecessary to strengthen the soft ground to avoid any failure
which may occur due to dynamic load, such as earthquake
load. Three classes of failure can be identified: internal
instability of the embankment, instability of the foundationsoil and overall instability involving both the embankment
and the foundation [2], [3]. Sometimes, beside thedeformation and failure of the embankment, lateraldisplacement of the ground affects the surrounding structures
which are located in a close proximity to the construction site.
To prevent any damage to the surrounding structures or
failure of the embankment it is inevitably required reinforcing
the subsoil beneath the embankment. In recent years, many
designers around the world are specifying base reinforcement
as one of the solutions for the short-term instability to makeuse of the tensile strength of the reinforcement to limit the
spreading of the embankment and lateral displacement of thefoundation [6], [8]. In this research, effectiveness ofreinforcing in embankment ground has been investigated andreinforcing mechanism has been clarified by laboratorymodel tests and the corresponding numerical simulations. To
analyze the behavior of reinforced earth structure on soft
ground, it is necessary to consider – (i) the elastoplastic
behavior of soil, (ii) soil/reinforcement interaction, and (iii)
soft ground consolidation systematically and simultaneously[1]. The numerical analyses are carried out with a finite
element program called FEMtij-2D, using the elastoplastic

Fig. 1 Apparatus for laboratory model
Tablet I:tPatterns of experiments

71

of the sheet pile from the edge of the loading plate, Hw, is 3cm.
In the nailing method, a tracing paper is being spread into theground with an angle of 30° from the horizontal direction. In
the tracing paper, aluminum rods were glued with an intervalof 1cm to provide frictional behavior in the nailing. The load
is applied on the loading plate with a motor which is attachedin the loading device, and the magnitude of the load ismeasured with a load cell installed in the tip of the device. A
slider is attached in the loading device to permit the lateraldisplacement of the base loading plate. Photographs are taken
during the experiments and they are used later as input data
for the determination of ground movements with a program
based on the technique of Particle Image Velocimetry (PIV).
Several model tests have been conducted varying the lengthof the sheet pile L=12cm, 24cm, thickness of the sheet pilet=0.3mm, 0.5mm, and the length of the nailing LN=0cm, 6cm,
9cm, to investigate the effects the stiffness of the sheet pile,and the lengths of the sheet piles and nailing into the grounddeformation. Among them six cases of experiments will be
discussed in this paper, the patterns are listed in Table I.
2.2 Numerical Simulations
An elastoplastic constitutive model for soils, called thesubloading tij-model [5], was used in finite element analyses.
This model requires only a few unified material parameters,
but can describe properly the following typical characteristics
of soils: (1) influence of intermediate principal stress on the
deformation and strength of soil; (2) influence of stress pathon the direction of plastic flow is considered by splitting theplastic strain increment into two components; (3) influence ofdensity and/or confining pressure. Model parameters for thealuminum rod mass are shown in Table II. The parameters arefundamentally the same as those of the Cam clay model [7]except for the parameter a, which is responsible for the
Table II. Material parameters for aluminum rods

0.008

0.004
N (eNC at p=98kPa
Same parameters as Cam- clay
0.3
& q=0kPa)
model
RCS=(1/3)CS(comp.) 1.80
e
0.20
of yield surface ( same as
1.20 Shapeoriginal Cam-clay at =1)a

1300

influence of density and confining pressure. Where,  and 
are the slope of loading and unloading curve of e-lnp graphs
at the loosest state. N is the void ratio at mean principal
stresses (p) 98kPa in the above mentioned loading curve ande is the Poisson’s ratio. The parameter  controls the shape ofyield surface. These parameters can easily be obtained fromtraditional laboratory tests. The parameters can be determinedthrough conventional triaxial tests and consolidation test. Fig.
2 shows the results of biaxial tests for the mass of aluminumrods used in the model tests. The figure shows the positive
and negative dilatancy of aluminum rod mass; and it is clearthat the strength and deformation behavior are very similar to
those of dense sand.
Fig. 3 shows the mesh used in the finite element analyses forthe analyses of the model tests. Isoparametric 4-noded
elements are used to represent the soil. The mesh is wellrefined with elements of finer mesh in most regions. The
sheet piles and soil nailing are modelled using elastic beam
elements. The frictional behavior (friction angle =18o)
between the reinforcements and the ground is simulated usingelastoplastic joint elements [4]. The frictional angle, =18o,
was obtained from a laboratory model test. Both vertical sidesof the mesh are free in the vertical direction, and the bottom
face is kept fixed. The analyses were carried out under planestrain conditions, since the aluminium rods do not deform in
the out of plane direction. The analyses are carried out withthe same conditions of the model tests. The initial stresses of
the ground are calculated by applying the body forces due toself-weight (=20.4kN/m3), starting from a negligible
confining pressure (p0=9.810-6kPa) and an initial void ratio
e=0.35. After self-weight consolidation the void ratio of theground was 0.28 at the bottom and 0.30 at the top.

Influence of density andconfining pressure

Fig.3 Mesh for Finite Element Analyses
3. RESULTS AND DISCUSSIONS
Fig. 4 shows the observed and computed results ofload-displacement curves. Figure (a) represents the results ofthe model tests, and figure (b) illustrates the results of thenumerical simulations corresponding to the model tests. The
vertical axis indicates the vertical load qv normalized with B(: unit weight of soil, B: width of the foundation) that
corresponds to the coefficient of bearing capacity, and thehorizontal axis is the vertical displacement normalized withfoundation width B. The observed load- displacement

Fig. 2 Stress-strain-dilatancy relation for the mass ofaluminum rods72

relations do not start form the origin, because of having theself-weight of the foundation. Comparing the results of case 2(sheet pile with L=24 and t=0.3mm) and case 3 (sheet pilewith L=24 and t=0.5mm) in the model tests, it is found thatthe stiffer and longer sheet pile produces a much stronger
reinforcing effect as generally expected. However, both case
1 and case 2 do not have a big difference with the result ofcase 1 (no sheet pile i.e. no reinforcement) when comparing itin the point of the ultimate bearing capacity. Hence, it can beunderstood that reinforcing with sheet pile alone does nothave much effect of reinforcement to increase the ultimate
bearing capacity of the ground if the ground is medium dense
soil. On the other hand, three cases of experiments where
nailing is combined with the sheet pile demonstrate a highbearing capacity including the ultimate bearing capacity inthe same ground. Comparing the results of case 4 (sheet pilewith nailing, L=24cm and t=0.3mm) and case 5 (stiffer sheet
pile with nailing, L=24cm and t=0.5mm), it is revealed thatthe bearing capacity is higher in case 5 until it reaches thefailure state. However, case 6, the same dimensions as case 5
except the length of the sheet pile (L=12cm), produces almost
equal amount of bearing capacity as case 6. It means that evena shorter sheet pile increases bearing capacity of the groundsignificantly if the sheet pile possesses a sufficient stiffness

and combined with a nailing. The computed results not only
qualitatively but also quantitatively predict well the observed
load-displacement curves of various kinds of tests patterns.
Fig. 5 illustrates the distribution of shear strain of the model
tests and finite element analyses during ultimate bearing
capacity. In the model test it is taken at v/B=0.06, and for thesimulation it is from v/B=0.10. The distributions of shear
strain of the model tests are obtained using Particle Image
Velocimetry (PIV) technique. In this paper, two images are
divided into a finite area; the average movement rate of themass of aluminum rods of each area is extracted as nodal
displacements. The strain for one grid is calculated from these
displacements by using the shape functions and the B matrix(strain- displacement matrix) that is usually used in finite
element method to relate displacements and strains. It is seenin the figures that for case 2 and case 3, where sheet pile alone
is employed for reinforcing, though it seems that the sheet
piles slightly restrict strain compare to the case of no sheet
pile (case 1) but larger shear strains concentrate locally at a
narrow zone in the shallow ground. Therefore, a significant
effect of reinforcement can no longer be achieved when sheet
pile alone is used as reinforcement of the ground. Moreover,
the reinforcing effect is not reflected in the bearing capacity

20

20

2qv/B

2qv/B(a) observed

15

10

10

0
0

(b) computed

Case 1Case 2
Case 3
Case 4
Case 5
Case 60.05

0.1

Case 1Case 2
Case 3
Case 4
Case 5
Case 6

5

v/B

0.15

0
0

0.05

0.1

Fig. 4 Vertical load vs. vertical displacement under concentric loadings

Fig. 5 Shear strain distributions

73

v/B

0.15

depth (cm)though the sheet pile is arranged in this way so that it canintercept the shear surface. It can be speculated that this isbecause of the upper part of the sheet pile which is displaced
to the opposite side of the loading plate. For case 4, case 5 andcase 6 where sheet piles are combined with nailing, it isrevealed that a wider region of shear strain with greater
intensity is developed even in the surrounding soil comparedwith case 1 (no reinforcement). Therefore, it can be said thatthe sheet pile where the nailing is combined can decrease theinfluence on the surrounding soil even if the sheet pile isshorter with higher stiffness. The shear strain of the numericalanalyses shows very good agreement with the results of themodel tests.
Fig.6 shows the lateral displacement of the left sheet pile atthe ultimate bearing capacity. It is seen that the lateraldisplacement of the upper part of the sheet pile is restricted in
the case when nailing is combined with the sheet pile. On theother hand, there is almost no difference of the lateraldisplacement at the upper part of the ground between the noreinforcing and the reinforcing with the sheet pile alone. The
reinforcing with the sheet pile alone does not produce asignificant effect of reinforcement which is confirmed in thisfigure. It is thought that the resistance force controls thelateral displacement because the tensile force acted on thenailing with the subsidence of the loading plate. Therefore, it
can be said that it is important to control the upper part of the
sheet pile to achieve a significant reinforcing effect.
0

the embankment with an inclination of 1:2. Fig. 8 shows themesh for the finite element analyses. Considering thesymmetric ground half section was taken for the analyses.
The width of the ground is 62.5m having 60m in depth. The
bottom face is assumed as fixed boundary condition. The
vertical faces are kept free in lateral directions. The analyses
are carried out considering soil-water coupled and planestrain conditions. The top surface of the ground is allowed todrain, and all other faces are assumed as impermeableboundaries. Therefore, water can not flow across theboundaries of the ground except the top surface. The watertable is assumed at the top of the ground. To consider the soft
soil, parameters of Fujinomori clay (bulk unit weight,
t=18.52kN/m³) is used as the base ground, and Toyoura sand(bulk unit weight, t=15.48kN/m³) is used as the material ofsoil fill. The material parameters are shown in Table III. The
coefficient of permeability for the ground is assumed as
10-7m/min. Two types of grounds, normally consolidated clay
(OCR=1.0) and over consolidated clay (OCR=2.0), are used
to investigate the dependency of ground stiffness. Fig. 9shows the stress-strain-dilatancy relation at triaxial condition
for (a) Fujinomori clay with normally consolidated conditionand (b) Toyoura sand with relative density Dr=75% underconstant cell pressure. The stress level chosen for the curve isequal to 294kPa which is the same vertical stress level at the
middle of the base ground considered in the finite element
analyses. The figures confirm that clay shows negative
dilatancy, while sand exhibits positive dilatancy. The
construction of the soil fill was achieved by gradually raising
the unit weight of the soil, and the construction is done inthree layers from the bottom to the top. After completion of
the construction, the analyses keep running until the complete

481216 (a)observed
20
24

Case 1Case 3
Case 5(b)computed
Case 1Case 3
Case 5-0.6 -0.4 -0.2 0 0.2 -0.6 -0.4 -0.2 0
lateral displacement (cm)Fig. 6 Lateral displacement of sheet pile

0.2

Fig. 7 Dimensions and ground types of embankment

4. SOIL-WATER COUPLING ANALYSIS OF REALEMBANKMENT

drained boundary
undrained boundary

In the previous section it was seen that the numerical analysis
can well reproduce the results of the model tests in the allcases of test patterns. In this section, soil-water coupling
analysis considering the construction process of real groundembankment will be discussed. In the introduction, it was
mentioned that embankments that are built on soft soils are
subjected to huge deformation or sometimes to failure of thewhole structure, which emphasizes the necessity ofstrengthening the soft ground to avoid any failure. Fig. 7
illustrates the ground types with the dimensions of theembankment to be analyzed in this section. The levee crownwidth is 5m, the bottom width is 25m, and the height is 5m of

Sheet pile

62.5m

Fig. 8 Mesh for FEA – real ground74

consolidation ratio OCR=1.0, the ground upheaves by about
2m immediately after completion of the soil fill in the case of
no reinforced ground, and it becomes about 1.5m after
complete consolidation at the toe of the embankment. For thesame ground, the subsidence of the embankment in thecentral portion of the base ground is about 2m immediately
after completion of the soil fill, and it is around 3m after
complete consolidation. In the reinforced ground with sheet
pile, the amount of the subsidence is controlled due to thereinforcement, for the stiff sheet pile the subsidence is about
0.6m immediately after construction and it is about 1.6m after
total consolidation in the central portion, for the flexible sheet
pile the embankment settles at around 0.8m after the-10
construction and about 1.8m after complete consolidation.
For over consolidation ratio OCR=2.0, the maximum
subsidence of the embankment is considerably less compare-5 to the normally consolidated soil (OCR=1.0). At the toe of theembankment, the ground upheaves immediately up to 0.35m
after the completion of the soil fill in the case of no reinforced
ground, and it becomes about 0.15m after complete
0
consolidation. For the same ground, the subsidence of theembankment in the central portion is about 0.4m

Table III. Parameters for Fujinomori Clay and Toyoura Sand
Fujinomori
Parameters
Toyoura SandClay
0.103900.070
0.009900.0045N (eNC at p=98kPa &0.92201.10
q=0kPa)RCS=(1/3)CS(comp.)
3.20
3.20
e
0.20
0.20

1.50
2.00
a(AF)=30a500a(IC)=500
Comp.

p=const.

1.5

2.5

Comp.

p=const.

q/p

2

1

1.5

(a) clay

1

0.5

d(%)
0

5

10

10

15

0.5

(b) sand

20 0

observed
calculated

0.5

5

v(%)

v(%)

2

q/p

observed
calculated

0

5

10

15

d(%)

20

Fig. 9 Stress-strain relation of clay and sanddissipation of excess pore water pressure which is generated
during the construction of the embankment. Similar to themodel test sheet piles with 2 types of stiffness having thelengths of 15m and 30m are used for reinforcing the ground.
In the case of composite reinforcing, nailing is added with the
sheet pile the same way as the model tests. The analysis
without reinforcing the ground has also been carried out tograsp the reinforcing effect in soft ground.
Fig. 10 represents displacement vectors of ground with
OCR=1.0 and OCR=2.0. The figures shows that the shape ofthe embankment is not maintained due to the excessive
deformation of the base ground having OCR=1.0 whenreinforcement was not installed (no reinforcement). In thiscase, the base ground flows outwardly during soil fill. It canbe seen that for the same type of the ground the upheaval of
the ground level is curbed in the reinforced ground with sheet
piles and composite reinforcement. It is because thereinforcement restricts the lateral flow of the base ground.
Comparing the results between the flexible and stiff sheet
piles, it is seen that larger displacement occurs in the sheet
piles with less stiffness for the ground of OCR=1.0. On theother hand, for the over consolidation ratio equals to 2.0, theshape of the embankment is maintained even in the casewhere the reinforcement is not installed. Therefore, in stiffer
soil reinforcement is not required in stiffer soil which
confirms the results of the other researchers.
Figs. 11 shows surface settlement profiles immediately after
construction and after complete consolidation. Here, thecomplete consolidation means that stage when the excess
pore water pressure completely dissipates.
For over(a) OCR=1.0(b) OCR=2.0

L=1B(15m),stiff
OCR=1.0

L=1B(15m),flexible
OCR=1.0

L=1B(15m),stiff
OCR=2.0(c) Composite reinforcement (OCR=1.0 & 2.0)
Fig. 10 Displacement vector in the ground2
1
0

no sheet pile
just after soil fillafter consolidation
with sheet pile
just after soil fillafter consolidation

-1
-2
-3
3

2
1
0

L=1B;15m; stiff
OCR=2.020

16-1

OCR=1.0
L=1B;15m; stiff
composite reinforcement
12

8

4

0

distance from centerline (m)

20

16

12

-2
8

-3
0

distance from centerline (m)Fig. 11 Surface settlement due to embankment
75

4

surface settlement (m)L=1B;15m; flexible
OCR=1.0

surface settlement (m)3

L=1B;15m; stiff
OCR=1.0

immediately after completion of the soil fill, and it is around
0.9m after complete consolidation. For the same groundregardless of the rigidity of sheet piles, the subsidence of theembankment in the central portion of the base ground is about
0.24m immediately after completion of the soil fill, and it isaround 0.76m after complete consolidation. However, it ismuch smaller in the case of the composite reinforcement thanthe other cases for both stiff and flexible reinforcing.
Fig. 12 shows the lateral displacement along the depths which
are located at the toe and 3m away from the toe of theembankment. Along the vertical direction from the toe of theembankment, the maximum lateral displacement occurs at the
point of 2m in depth from the ground level in the noreinforcement ground in the normally consolidation(OCR=1.0) ground, and the magnitude of the lateraldisplacement is 2.5m. Along the vertical line of 3m awayfrom the toe of the embankment, the maximum lateraldisplacement, about 2m, occurs at the surface and it decreases
with the increase of depth. Though the lateral displacements
along the line of 3m away from the toe of the embankment aresmaller than that of the line at the toe of the embankment, but
at the surface almost same displacement occurs. In thereinforcing ground with sheet pile alone, the maximum lateraldisplacement occurs at the surface for both locations. In thisground, the lateral displacement is about 0.12m in the case of
stiffer sheet pile, and it is about 0.35m for the flexible sheet
pile after complete consolidation. For the composite
reinforcement, the lateral displacement is smaller than theother cases. The lateral displacement along the line of 3maway form the toe decreases with the progress of
consolidation. As the excess pore water pressure dissipates
after the construction, the phenomenon of reducing lateraldisplacement can be speculated.
In the overconsolidated ground (OCR=2.0), along the verticaldirection from the toe of the embankment, the maximum
lateral displacement occurs at the point of 1m in depth fromthe ground level in the no reinforcing ground. Along thevertical line of 3m away from the toe of the embankment, themaximum lateral displacement (0.2m) occurs at the surface.
The lateral displacement decreases with the advancement of
the consolidation at both faces. In general, the magnitude oflateral displacement is smaller in the case of the stiffer sheet
pile than that of the flexible sheet pile the same as thenormally consolidated ground. As a whole, the reinforcing
with the sheet piles and soil nailing restrict the lateraldisplacement and the maximum displacement occurs at thesurface the same as the normally consolidated ground.
5. CONCLUSION
In this research effectiveness of reinforcing in embankment
ground has been investigated with laboratory model tests andnumerical simulations. It is found that in the case of thereinforcing with the sheet pile alone, no reasonable increaseof the bearing capacity of the reinforced ground is observed
as the upper part of the sheet pile is displaced outwardly withthe loading even when it is set up in the position where thewidening of the shear surface is restricted well. In contrast, inthe case of the composite reinforcing method (sheet pile withnailing), a significant increase of the bearing capacity is seenas the lateral displacement of upper part of the sheet pile isimpeded due to the inclusion of the nailing. Even the shorter
length of the sheet pile produces almost the same effects as
those of the longer sheet pile when nailing is appended in thereinforcing method. The finite element program FEMtij-2D
properly predicts the results of the model tests on thereinforced embankment. The subsidence of the embankment
and the lateral displacement of the ground can be reduced
introducing reinforcement in a real ground. The stiffer sheet
pile and the composite reinforcement where soil nailing is
combined with the sheet pile cause a significant reinforcing
effect in a soft ground. The finite element analyses can give a
guideline for the prediction of deformation pattern and for the
optimum dimensions of the reinforcement in the soft ground.
REFERENCES
[1]

0

[2]
5

3m outside from toe
just after soil fillafter consolidation

depth (m)no sheet pile
OCR=1
toe of soil fill
just after soil fillafter consolidation[3]

10
Sheet pile
L=1B;15m; stiff
OCR=1

Sheet pile
L=1B;15m; flexible
OCR=1

[4]
15
0

[5]
5

Sheet pile
L=1B;15m; stiff
OCR=2
-2
-1.5
-1
-0.5
0
-0.2 -0.15 -0.1 -0.05 0
lateral displacement (m)lateral displacement (m)depth (m)no sheet pile
OCR=2

Composite
reinforcement
L=1B;15m; stiff
OCR=1-0.2 -0.15 -0.1 -0.05 0
lateral displacement (m)

[6]

10

[7]

15

[8]

Fig. 12 Lateral displacement at two sections along depth76

Bergado D. T, Chai, J. C. & Miura N., “FE analysis of grid reinforced
embankment system on soft Bangkok clay”, Computers andGeotechnics, Vol. 17, No. 4, 1995, pp. 447-471.
Hird C. C. & Jewell R. A., “Theory of reinforced embankments”,
Reinforced Embankments: Theory and Practice in the British Isles,
London: Thomas Telford, 1990, pp. 115-139.
Hird C. C., Pyrah I. C., & Russell D., “Finite element analysis of thecollapse of reinforced embankments on soft ground”, Geotechnique,Vol. 40, No. 4, 1990, pp. 633-640.
Nakai T., “Finite element computations for active and passive earthpressure problems of retaining problems”, Soils and Foundations, Vol.
25, No.3, 1985, pp. 98-112.
Naka T. & Hinokio T., “A simple elastoplastic model for normally andover consolidated soils with unified material parameters”, Soils andFoundations, Vol. 44, No.2, 2004, pp. 53-70.
Monnet J., Galera I. & Mommessin M., “Some theoretical approaches
about reinforced embankments on weak soil”, Computers andGeotechnics, Vol. 7, No. 1, 1989, pp. 37-52.
Roscoe K. H. & Burland J. B., “On the generalized stress-strain
behaviour of ‘wet’ clay” J. Heyman & F. A. Leckie (eds.), Engineering
plasticity (Cambridge: Cambridge University Press), 1968, pp.535-609
Sharma J. S. & Bolton M. D., “Finite Element Analysis of Centrifuge
Tests on Reinforced Embankments on Soft Clay”, Computers andGeotechnics, Vol. 19, No. 1, 1996, pp. 1-22.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Evaluation of Stamps and Latex Shell Influence on Stabilometer Testing
Results, Modeled with Simulia Abaqus
Dr. Dmitriy Volik, Kuban State Agricultural University, Russian Federationa model that describes the system behavior: stamp – soil
sample – latex shell. If necessary, parameters are calibrated;graphs, obtained on their basis, are compared with laboratory
data. Therefore, it appears possible to obtain a similar
behavior of a mathematic model and a triaxial machine.
However, the above mentioned parameters simulate thebehavior of the “stamps – soil sample – latex shell” a system,
whereas parameters, which simulate the behavior of the soil
sample only, are required in case of geotechnical tasks. The
main goal is to obtain parameters of the soil sample only,
considering that an “ideal” triaxial machine, which would test
the soil sample without stamps and latex shell influence, does
not exist.
This work set outs an analysis of stamps and latex shellinfluence on the results of soil tests in a triaxial machine. In
order to evaluate the influence (to allow for an inaccuracy) of
stamps and latex shell in the triaxial machine, a modeling of athree-dimensional “stamps – soil sample – latex shell” system
is carried out through the program complex of Simulia
Abaqus, following a 4-stage methodology:
The first stage involves creation of a three-dimensional
system model (stamps – soil sample – latex shell) in the
Simulia Abaqus program complex. It will be referred to as astabilometer model further on.
The second stage includes verification of stabilometer model
parameters to conform with the behavior of the triaxial
machine. Verification has to be carried out for each soil
sample separately. In the given example it is middle-size
grain sand and loamy clay.
The third stage requires exclusion of the stamp and latex shellfrom the stabilometer model, thereby obtaining an “ideal”
stabilometer model.
The final stage involves a comparative analysis of the
behavior of the stabilometer model and the“ideal”
stabilometer model, in order to evaluate the influence of thestamp and latex shell.

ABSTRACT: Triaxial tests is often used, along with other researchmethods, in order to define parameters of geotechnical data. The
main topic being conferred in work is analysis of influence, exerted
by stamp and latex shell, upon the results of triaxial machine soil
testings. The stabilometer modeling and verification was produced.
Calibration of soil model parameters, derived from triaxial tests, isbrought in work. The necessity of three dimensional simulation ofsoil testing process is pointed out in order to take into account stampand shell influence.

Keywords: Triaxial machine test, soil model, Simulia Abaqus
modeling, stamp and shell influence, modified Drucker–Prager/cap
plasticity model.

1. INTRODUCTION
Nonlinear models are frequently used in geotechnicalcomputational
mechanics,
which
allow
variousimplementation features of soil behavior to be taken into
account. In order to define parameters of nonlinear soilmodels, data of triaxial tests is often used, along with otherresearch methods.

Fig. 1. Triaxial machine (NPO «Geotek»), and its principal
scheme

2. STABILOMETER MODELING WITH SIMULIAABAQUS

As generally known, uniform hydrostatic pressure σ2=σ3 ismaintained in the triaxial machine camera. In order to divide
mediums of stabilometer camera and soil pores the sample iscontained inside a flexible latex shell, while the verticalcomponent of deviatoric load σ1 is imposed on the soil sample
via firm stamps at the bottom and at the top (Fig. 1.). As aresult, during the experiments the soil sample interacts withthe stamps and the latex shell via their contact surfaces.
Therefore, based on the triaxial tests we obtain parameters of

1/8 part of stabilimeter (stamp – soil sample – latex shell) was
simulated in Simulia Abaqus. As a visualization of the results,
1/8 part was reflected/ copied in order to receive a visual
effect of the model second part presence (Fig. 4.)
2.2 Soil model
In order to describe a soil behavior, modified
Drucker-Prager/Cap model is used in the analysis.
The cap model is appropriate to soil behavior because it is77On the second stage, vertical pressure σ1 is increased with amaintained compression σ2 (σ3) till a relative vertical strain
achieves ε1=12% [1].

capable of considering the effect of stress history, stress path,
dilatancy, and the effect of the intermediate principal stress.
The yield surface of the modified Drucker – Prager/cap
plasticity model consists of three parts: a Drucker–Prager
shear failure surface, an elliptical Cap, and a smoothtransition region between the shear failure surface and the
Cap (Fig. 3.).

Fig. 3. Drucker–Prager/cap plasticity model yield surface
The Drucker–Prager shear surface is a perfectly plastic yield
without hardening. Plastic flow on this surface produces
inelastic volume increase - dilation.
The Cap surface bounds the yield surface in hydrostatic
compression. On the Cap surface plastic flow causes thematerial to compact correspond to hardening mechanism.
Besides, the Cap surface helps control volume dilatancy whenthe material yields in shear by providing softening as afunction of the inelastic volume increase created as thematerial yields on the Drucker–Prager shear failure andtransition yield surfaces. The model uses associated flow in
the Cap region and nonassociated flow in the Drucker–Prager
shear failure and transition region.
The hardening/softening law is a user –defined piecewise
linear function relating the hydrostatic compression yieldstress, and the corresponding volumetric plastic strain.
The nonlinear elastic behavior modeled by using the porous
elasticity model including tensile strength. During
compression, the void ratio decreases. This behavior isexpressed by the logarithmic bulk modulus.

Fig. 4. Finite elements mesh – 1/8 parts, more than 900 nodes.
Deformed model - second part (reflection/ copy of 1/8 parts
relative to XY and YZ)
During the triaxial tests, soil samples, which are used, arethose of moist middle-size grain sand and loamy clay, of 38
mm in diameter and 76 mm in height.
The latex shell thickness is 0,1mm.
3. STABILOMETER MODEL VERIFICATIONFirst of all, it should be mentioned, that characteristics ofground model are standardized for a full compliance of a its
model behavior with the results of laboratory tests according
to the relation ε1(σ1). Calibration is carried out for several
curves with a different value of the uniform compression σ2(σ3). Example of the compliance of stabilometer model
behavior with the results of triaxial tests of one of the curves
is illustrated on the (Fig.5).
For a verification, deflected mode of stabilometer model is
examined and compared to the features of behavior oflaboratory ground samples in the same conditions.
Analysis of a deflected mode of the stabilometer model isgiven below.
Following features, relevant for a given process, are shown in
the Fig. 5:
1. Maximum magnitudes of stress intensity in soil arelocated in the shape of a “cross”, which is registered oncomputed tomography using of full-scale specimen [3].
2. Relative sliding and local break away of a latex shellfrom a stamp and soil.
3. Latex shell, consequently, acquires a barrel-type
shape [2].
4. Soil flowing under the latex shell [2].
5. Stress intensity has a similar distribution to both

2.3 Stabilometer modeling
Stamps and shell are modeled within the limits of linear
elasticity. Steel and rubber has been accepted as materials forstamp and latex shell respectively. Non-linear contact
between a stamp, soil sample and latex shell is simulated.
Model was meshed by C3D8 solid elements.
For soil initial conditions production the Geostatic Abaqus
procedure was applied.
Load is imposed to a model in two stages, repeating
laboratory experiments in stabilometer, according to the CD
scheme [1]:The first stage involves sample recompression via a uniform
hydrostatic pressure σ2 (σ3) to the natural level of a proper soil
weight σν. This stage is a very important because this will
determine the initial stresses in all soil elements.
78

Behavioral curves of a sandy soil register an insignificant
influence of stamps and shell in the stabilometer model. This
type of a soil is often used as a base for pile foundations in thecity of Krasnodar.
Stamps and shell influence in the system with a clay soil may
comprises up to 6%. Therewith, a loamy clay is used as a basefor building foundation slabs in the constructions in
Krasnodar, reaching 16-17 floors.
It is necessary to compare plastic strain in the stabilometermodel and the“ideal” stabilometer model, in order to evaluate
the influence of the stamp and latex shell.

models made in Abaqus and LS-DYNA [2] programs.

Fig. 5. Stress intensity according to Mises criterion, Pa.
Left fig. – Simulia Abaqus; Right fig. – LS-DYNA [2]A fact of presence of listed features in tests as well is an basis
for a deduction about the qualitative similarity in a behaviorof stabilometer model.
Fig. 7. Equivalent plastic strain at integration points.
Left fig. – the stabilometer model; Right fig. – the “ideal”
stabilometer model

4. THE STAMP AND LATEX SHELL INFLUENCEEVALUATING
Specific differences in a behavior of stabilometer model and amodel of an “ideal” stabilometer are represented on thegraphs below Fig. 6.

Apparently from the Fig. 7, ground samples undergo various
deformed states. Ground in a stabilometer model has abarrel-shaped strained shape (presumably, due to a shellinfluence), whereas ground sample in the “ideal” stabilometerhas a regular cylindric shape.
Moreover, intensity of plastic strains in stabilometer model
are placed in a shape of a cross.
In a model of the “ideal” stabilometer, intensity of plasticstrains in distributed equally in the whole sample. Tosummarize, it can be resumed that the stamps and shells
influence not only in quantity, but also in quality.
Difficulties occur in evaluating the influence of stamps andshells, relying only on the difference of 6% between curves
on a graph point, which is close to the critical state of asample. It may be assumed that, in some spots under thefoundation, conditions are far from the limit equilibrium of aground.
However, a practical task is suggested to be solved, in order toevaluate an influence of stamps and shell.
Thereto, calculation of a tank settlements on a loamy clay was
carried out. Calculation was carried subject to a groundconsolidation during hydrotests of a tank and next several
cycles of a filling-drawdown.
A tank with a diameter of 11 meters, product lever 22meters.
Ground base with a calculated zone 20 meters deep.
Permeability coefficient of a loamy clay k=0,000001 m/c.

Fig. 6. Relation graph ε1(σ1)
As is obvious from the graphs, the sample destruction occurs
significantly earlier without a stabilizing influence of a latex
shell and a stamp.
79

Calculations are performed with parameters of foundation
grounds, defined including stamp and shell influence, andwithout it.

behavior of a sample, shell and a stamp. Consequently, model
parameters, based on these curves, which are used forpredictions of a soft soil behavior, should be applied withadjustments.
In order to detect parameters of a soil model, it is necessary totake into account stamp and shell influence by means of
three-dimensional simulation of a sample testing process.
6. REFERENCES
[1]
[2]
[3]

Fig. 8. Vertical displacements in models, m.
Left side – adjusted soil model data; Right side – non adjusted
As shown at the Fig. 8., tank foundation settlements are
significantly underestimated with unadjusted soil model
parameters, that have 0,10m. Maximum base settlement
under reservoir foundation with adjusted soil model data is
0,11m. The underestimated settlement is 9% even with thecoarse elements mesh model.
5. CONCLUSION
Via Simulia Abaqus program complex, a system model“stamps – soil sample – latex shell”, has been simulated,which makes it possible to predict laboratory tests in thetriaxial machine.
Stabilizing influence of shell and stamps enables a soil
sample to describe a curve ε1 (σ1) of a twice greater length till
the moment of destruction.
A shell and stamp contribution considerably increases a
system stiffness “stamps – soil sample – latex shell”, in caseof soft soil tests (clay soil of plastic consistency, loose finesand etc.).
Ground in a stabilometer model has a barrel-shaped strained
shape (presumably, due to a shell influence), whereas groundsample in the “ideal” stabilometer has a regular cylindricshape. Intensity of plastic strains in stabilometer model areplaced in a shape of a cross. In a model of the “ideal”
stabilometer, intensity of plastic strains in distributed equallyin the whole sample. To summarize, it can be resumed that thestamps and shells influence not only in quantity, but also in
quality.
The underestimated maximum settlement of the tankfoundation on a loamy clay base is 9% with unadjusted soilmodel parameters.
Stabilometric experimental curves represent a combined
80

GOST 12248-96. Soils. Methodology of laboratory assessment ofstrength and deformability properties. Russian Federation standard.
Moscow. – 1997.
Muiseimnik A. Y. Identification of soil model parameters / A. Y.
Muiseimnik, G. G. Boldyrev, D. V. Arefyev // Engineering Geology. –2010. - №3. – p. 38-43.
Batiste, S. N. Shear Band Characterization of Triaxial Sand Specimens
Using Computed Tomography / S. N. Batiste, K. A. Alshibli, S. Sture,
M. Lankton // Geotechnical Testing Journal/ -Vol. 27. - -№6. – 2004.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Characterizing the Interaction of Geomat with Fine Grained Black Cotton
Soils
Wekesa S., Mukabi J. & Okado J., Research &Development Department, Kensetsu Kaihatsu Ltd, Nairobi, Kenya
Kogi S. & Ndeda M., Materials Testing & Research Department, Ministry of Roads, Nairobi, Kenya
Ogallo J., Department of Civil Engineering, University of Nairobi, Nairobi, Kenya



ABSTRACT: Black Cotton Soils (BCS), which are highly
problematic in nature when it comes to their use as construction
materials for geo-structures, are normally in extensive existence inmost parts of Africa. As a consequence, it is extremely important toadvance Research & Development (R & D) related to BCS in orderto develop cost-effective geotechnical engineering methods of its
utilization. On the other hand, over the past 20years, tremendous
advances have been made in the research and development ofgeosynthetics technologies and their appropriate geotechnicalengineering applications.
Results based on laboratory tests are reported and the role of theincreased interaction between the geomat and BCS in the functional
aspects is quantitatively analyzed through the application ofadvanced analytical tools. It is derived and concluded that the degree
and extent of the geomat-BCS particle interaction is dependent on
several factors mainly including particle size, texture, shape,moisture-suction variation and location of embedded geomat
relative to the layer thickness.

Keywords: BCS, Soil particle-geomat
Characteristics, Geosynthetics.

interaction,

1. INTRODUCTION
Black Cotton Soils (BCS) are usually tropical clays which areessentially products of physical and chemical in-situ
weathering of igneous, sedimentary and metamorphic rocks
under varying environmental conditions. The formation ofthese soils is highly influenced by a complex interaction ofvarious variables such as weathering, erosion and climaticchanges as well as type of original parent rock, local
topography, drainage and cycle factors which strongly
influence their geotechnical engineering behavior. As aconsequence, their characterization for use as suitableengineering materials is exceedingly complex since they arehighly susceptible to various environmental changes.
Given the recent advances that have been, and are being madein the research of geosynthentics materials based onmonitoring of the instrumented structures throughout theyears, new design methods have been realized and theversatility of their applications for soil reinforcement,
environmental protection and performance during major
earthquakes, have increasingly become appreciated [1]-[4].
Nevertheless, hardly any research has been devoted to an
in-depth study of the quantitative characterization of the finegrained soil particle-geomat interaction in terms ofenhancement of strength and deformation resistance [5], [6].
This study therefore, considers the importance of undertaking
comprehensive geo-scientific and geotechnical engineering

investigations into the fundamental aspects of the interactive
properties thereof.
Based on some of the typical results from this study, it isderived and concluded that the degree and extent of thegeomat-BCS particle interaction is, besides the intrinsic
properties of the BCS and also dependent on the layer
thickness defining the active zone and confining pressuresand tensile forces.

2. INTRODUCING SOME TYPICAL BCS PHYSICAL,CHEMICALANDMECHANICAL
CHARACTERISTICS
2.1 Typical Intrinsic Physical Properties and BasicClassification
Table 1 is a summary of the typical intrinsic physicalproperties of the BCS encountered in most parts of the East
and Central African Region [7]. The global classification of
BCS, on the other hand, is presented in Table 2 [8].
Table 1 Summary of typical physical properties of BCS
Property

Value

Liquid Limit LL (%)

37~83

Plasticity Index PI (%)

16~45

Moisture Content w (%)

16~40

Shrinkage Limit SL (%)

8~20

Percent Passing 75  m

75~95

Free Swell (%)

70~165Swelling Pressure kN/m2

0~80

Linear Shrinkage (%)

18~35

Specific Gravity Gs

2.47~2.56

Dry Density  d kg/m3

1920~2050

Table 2 Global classification of BCS
Soil Parameter

Classification
Moderate Swellability

High Swellability

Very high Swellability<15kN/m3

15<  d < 16

>16

Clay content C<0.002 <40%

40<c < 55

>55

Liquid Limit (LL)

<48%

40<LL < 55

>65

Plasticity Index (PI)<30%

30<PL < 40

>40

Shrinkage Index (IS)

<30%

30~60

>60

Swell Pressure<120kN/m2

120~600>600

Dry Density   d

It can basically be derived from these tables that these soils
are extremely fine grained and highly susceptible topore-grain moisture-suction matrices characterized mainly by
swelling due to wetting and shrinkage due to drying (alsorefer to Fig. 2.1 [8]-[11].
2.2 Response to Physical-chemical Variations andEnvironmental Changes
A study was carried out by [9]-[11] to determine the responseof BCS as influenced by varying physical-chemical andenvironmental conditions. Fig. 2.1 depicts part of the typicalresults from these reference citations.
To simulate the effects of rapid drying on the PI

Plasticity properties of Black Cotton Soils(b)

Phase Stabilisation
Drying

WettingDryingPI Value (%)

33
31
29
27
25
23
21
19
17
15

Wetting

Wetting

Rapid drying effects of PI

35
34
33
32
31
30
29
28
27
26
25
0

0

24
48
72
96
120
Wetting&Drying Period (Hrs)
Neat
Con‐Aid
Con‐Aid+5%Lime

8 10 12 14 16 18 20 22 24
Drying Period (Hrs)
Untreated (Air dried)
Con‐Aid treated (Air dried)
Untreated (Oven dried)
Con‐Aid treated (Oven dried)

144

Figure 2.1 Simulation of impact of drastic environmental
changes on black cotton soil subjected to varying
physical-chemical conditions
Furthermore, the characterization of BCS subjected tovarying drying/wetting (environmental changes) withdifferent chemical additives can be observed from Fig. 2.1.
These results basically show that BCS are highly influenced
by changes in environmental conditions and that the degree ofthese changes depends on their current physical-chemical
state. It can further be noted from Fig. 2.1 (b) that theintensity and period of the drying is significantly influenced
by the chemical additive. Detailed analyses of the interactive
characterization of this behavior are reported in [9].
2.3 Typical Shear Strength Characteristics when
Subjected to Moisture-Suction Variation
As stated in the introduction, the behavior of BCS is highly
influenced by the changes in moisture-suction conditions.
Fig. 2.2 shows that: 1) a slight increase in moisture content
can significantly impact on the strength magnitude andoverall characteristics of the BCS; and, 2) at higher moisture
contents, the effects on characteristics of dynamically loaded
and statically loaded BCS specimens are significantly
different in the initial and subsequent yield surfaces. However,at failure, the maximum strengths and residual behavior are
virtually the same. This may imply that, similar to most of the

2

4

6

other geomaterials, BCS are sensitive to various intrinsic andextrinsic influences within the initial, secondary and tertiary
yield surfaces [12]. This fact can also be confirmed from Figs.
2.4 and 2.5 in sub-section 2.4, which depict the effect of
Cyclic Prestraining (CP) under controlled loading conditions
whereby the bearing capacity, strength and deformation
resistance tend to be enhanced [13], [14].
From the foregoing perspectives, it was considered that the
study of the influence of geomat reinforcement on thegeotechnical engineering behavior of BCS may make aninteresting research subject.
2000
Axial Stress, UCS (kN/m2)

PI Value (%)(a)

characteristics of CON-AID treated material, a comparison of
air and oven (110°C) dried untreated (neat) and treatedmaterial was carried out. Figure 2.1(b) is a plot of theseresults. It can be noted that CON-AID treated material isneither influenced by rapid drying effects nor curing period
and maintain the same magnitude of PI (28) as in previous
cases.

Proposed by Mukabi et al. (2001c) for
Tropical Soils with a PI<43

Dynamic (Mc=10.2%)

q  1800

 mc u vmc  0.0123
PI0.535

1600

.1.9  Emax vmc  0.0362
PI1.9qu imc



mc

1400





Emaximc  E  mc max vmc   23.5CS CC  6.44
 Emaximc 
Cs = Swellingwhere,  mc = 0.53
Index, CC=Compression Index

.

1200
1000
800

Static (Mc = 12.1%)

600
400
Dynamic (Mc=12.1%)

200
0
0

0.3 0.6 0.9 1.2 1.5 1.8 2.1 2.4 2.7

3

3.3 3.6 3.9 4.2

Axial Strain, εa (%)

Figure 2.2 Effect of moisture-suction variation on statically[10].

Axial Stress, qu (kN/m2)

0

0.3

450
400
350
300
250
200
150
100
50
0
‐50

0.6

0.9

Axial Strain, εa (%)
1.2 1.5 1.8

2.1

2.4

2.7

3

Unsoaked
SoakedSr = Saturation Level

Sr = 71.8%

D

Regressional relation

Linear (Regressional relation)

6000
Moisture‐Suction Variationin relation to a Reference
MC of 10.1% for 2% Lime
Treated Black Cotton Soilunder UC Static Loading

5000
4000

Sr = 100%

3000
2000

Figure 2.3 Effect of saturation level on the shear strength ofblack cotton soil.
The results in Figs. 2.2 and 2.3 are a clear indication of thefact that BCS is extremely sensitive to the slightest ofmoisture-suction variation and that the degree of thisinfluence can be quite detrimental.
2.4 Typical Shear Strength and Deformation Resistance
Characteristics as a Result of Controlled Stage Loading
Figs. 2.4 and 2.5 show the characteristics of BCS subjected tocontrolled stage loading within the initial yield surface,constituent also of CP effects [12].

Trial1

1.8

Cu,(Kgf/cm2)

The results depicted in Figs. 2.4 and 2.5 are a clear indicationthat the shear strength and deformation resistance of BCS canbe enhanced significantly when subjected to controlled stage
loading within the Initial Yied Surface {(εa)ELS(YI)} [12], [14].

Resilient Modulus of BCS,Mr(Kg/cm2)

and dynamically loaded tropical black cotton soil.

Trial2

Trial3

Stage‐Loading Methods Designed to
Effectively Enhance Engineering
Properties of Expansive Soils

1.4
1

Role of Increase In Bearing Capacity
Dependent Pavement Structure ofUpper Layers i.e Surcharge Pressure

0.6
Post Construction
Static Loading.

0.2
0

1

Static Post
Dynamic Loading

Post Primary
Consolidation2
Loading Stages

3

4

Figure 2.4 Effect of long-term static and dynamic stage
loading on the shear strength properties of black cotton soils[10].
Resistance of Black Cotton Soil
10
9
8
7
6
5
4
3
2

(εa)ELS(%)X10‐4

Increased Loading within
Initial Yield Surface
Increases Deformation
Resistance01

Trial 1

2
Loading Stages

Trial 2

3

4

Trial 3Figure 2.5 Effect of long-term static and dynamic stage
loading on the deformation resistance of black cotton soils

1000

Mr = ‐497ΔWC + 5431

0
0

2

4
6
Water Content, ΔWC (%)

8

10

Figure 2.6 Influence of moisture-suction variation on theresilient modulus of slightly lime treated black cotton soil atequilibrium moisture content.
On the other hand, Fig. 2.6 shows the susceptibility of BCS tomoisture~suction variation. The following derivations can bemade from these figures: 1) the effects of moisture-suction
variation are quite steady once the BCS is treated with a 2%
hydrated lime (chemical) content; 2) a virtually linear relation
seems to exist between the resilient modulus with the increasein moisture content in this case. This may be implicit of thefact that the chemical treatment (the addition of hydratedlime) somehow contained and streamlined the aggregation
and agglomeration of the BCS particles. This particularinteraction of physical-chemical containment is one of thefeatures of correlative interest of this study in regard towhether or not the geomat reinforcement would exhibit
analogous and/or virtually similar effects in reference to theimprovement of the vital geotechnical engineering properties,
as do the chemical additives; and, 3) slight increases inmoisture content lead to a drastic reduction in the resilient
modulus of the BCS notwithstanding the control chemicaltreatment.
3 METHODS OF TESTING
Conventional and modified laboratory and field methods oftesting were adopted in this Study. The main mechanicalmethods of laboratory testing included the conventional andmodified Unconfined Compressive Strength (UCS) adopting
larger size specimens (φ ≥ 20cm), conventional Consolidated
Undrained Triaxial Compression (CUTC), Consolidated
Drained Triaxial Compression (CDTC), modified simple
shear and modified UCS dynamic loading under varying
moisture – suction conditions and modes of reinforcement.
In order to correct for the specimen size, a full scale plateloading test was carried out in the field. Details of themethods of testing are reported in [15].
4 DISCUSSIONS OF TEST RESULTS
4.1 Stress-strain Relations
Fig. 3.1 shows the stress-strain characteristics of theinteraction between BCS particles and geosynthetics. Of
particular interest is the behavior of geomat reinforced BCS.
The specimens were tested under varying environmental
conditions as well as modes and types of geosynthetic
reinforcement. All imbedding locations are measured fromthe bottom of the specimens.
Geosynthetics+Fine Grained Slightly Sandy BCS

4of OPMC stabilization, the contribution of geosynthetic
reinforcement is hardly noticeable in the compression zone.
In this case, it can therefore be concluded that whereas
geosynthetics contribute largely to mobilization anddistribution of tensile stresses and containment of tensile
straining OPMC contributes immensely to the enhancementof compressive stresses and control of compressive straining[5].

7

Effect of Various Geosynthetic Location and OPMC
Stabilization
Legend6

UCS (kgf/cm2)

5

TensarGeogrid,Neat,1dc 1ds

Stabilzed with Geosynthetic

Double Tensar
Mat,Stabilized,1dc
1ds
Tensar
Mat,Stabilized,1dc
1ds

3

3
2

1

Geogrid, 1/3H + Neat
BCS

1/3
1/5
1/3,2/3
Position of Geosynthetic in Relation to Specimen Height

Geogrid, 1/3H,2/3H +Neat BCS
Mat, 1/3H + Neat BCS

Figure 3.2 Effects of location and stabilization on thegeosynthetics-BCS particle interaction as a measure of UCS.

Mat, 1/5H + Neat BCS

0Mat, 1/3H,2/3H + Neat
BCS

0

1

2
3
4
Axial Strain, εa (%)

5

Neat without Geosynthetic0

Geogrid, 1/5H + Neat
BCS

1Neat with Geosynthetic

2

No Soil Reiforcement(Neat BCS)

6

Figure 3.1 Stress-strain characteristics of reinforced and neat
black cotton soil.
The following derivations can be made from Fig. 3.1: 1) thecontribution of both geomat and geogrid reinforcement to theenhancement of the Unconfined Compressive Strength (UCS)of BCS is highly influenced by the location where it isimbedded; 2) geomat reinforced BCS exhibits better andsuperior stress-strain characteristics when compared with thegeogrid reinforced BCS. This can mainly be attributed to thefact that, due to the mesh size of the geogrid relative to thegrain size of the BCS, the interlocking mechanism issignificantly limited curtailing the role of the geogrid as areinforcing element; 3) the double mat reinforcement, which
resulted in the highest strength and deformation resistanceseems to have a casing effect whereby stress mobilization anddistribution are simultaneously activated; and, 4) the
stress-strain and failure characteristics are influenced by themode and type of reinforcement in all cases.
In order to study the impact of varying modes and types ofreinforcement on the behavior of agglomerated BCS particles,
OPMCS (Optimum Mechanical and Chemical Stabilization)[16], was introduced into the testing regime. Fig. 3.2 is asummary of some of the typical UCS results that wereobtained from this study. Reference can also be made to Figs.
3.3 and 3.4. It can be noted in all cases that the optimum
location for single reinforcement for both geomat and geogrid
is 1/5 from the bottom of the specimen for non-stabilizedgeomaterials. It can further be observed that due to the effect

4.2 Angle of Internal Friction
The objective of measuring the angle of internal friction was
mainly to correlate the parameter to the interlocking
mechanism of the geosynthetic and geomaterial particles.
Some typical results are presented in Fig. 3.3.

30
Angle of Internal Resistance, φ (ͦ)

UCS, qu (kgf/cm2)

4

Effect of Various Geosynthetic Location and OPMC
Stabilization

25

LegendStabilized with Geosynthetic

TensarGeogrid,Neat,1dc 1ds

20

Double Tensar
Mat,Stabilized,1dc 1ds

15
10

Tensar
Mat,Stabilized,1dc 1ds

Neat without Geosynthetic

Tensar Mat Neat,0dc
0ds

5

TensarGeogrid,Neat,1dc 1ds

0
1/3

1/5

1/3,2/3

Position of Geosynthetic in Relation to Specimen Height

Figure 3.3 Effects of location and stabilization on thegeosynthetics-BCS particle interaction as a measure of angle
of internal resistance to simulate interlocking mechanisms.
The results in this figure confirm the findings reported in
section 3.1. However, it can be noted that: 1) at high moisture
contents, BCS exhibits very low internal resistancenotwithstanding geosynthetic reinforcement. Consequently,
moisture control in BCS is quite essential in all cases; 2)
OPMC stabilization contributes significantly to theenhancement of the magnitude of angle of internal friction

due mainly to particle agglomeration; and, 3) presumably due
to the monolithic state that it assumes after aggregation andagglomeration, the location, mode and type of reinforcement
have practically no influence on the OPMC stabilizedgeomaterials in this case.
4.3 Elastic Stiffness
The elastic stiffness was derived in terms of Young’s
modulus as depicted in Fig. 3.4. Similar to the cases reported
in the preceding sections, the influence of location, mode andtype of geosynthetic reinforcement on the magnitude of theelastic stiffness was investigated.
1450

Effect of Various Geosynthetic Location and OPMC
Stabilization

Figure 3.5 Example of typical shear banding and failure
characteristics.
Fig. 3.5 shows some typical shear banding and post-failurestate of OPMCS plus geomat reinforced specimens.
It can be inferred from this figure that: 1) significant shear
banding occurs within the interface of the geomat and thestabilized geomaterial; 2) the section that is sandwiched by
the geomats hardly exhibits any shear banding; 3) the
specimens exhibit fairly uniform failure characteristics; 4)due to scale effect, failure seems to have been constrained in
the compression zone thus the occurrence of failure at theinterfaces; 5) due to 4) above, it is imperative to further
investigate these characteristics on larger specimens andfull-scale experimental sections.

Young's Modulus, Emax (kgf/cm2)

1400
1350

Legend1300
1250

Double Tensar
Mat,Stabilized,1dc
1ds
Tensar
Mat,Stabilized,1dc
1ds
Tensar Mat Neat,0dc
0ds

Stabilized with Geosynthetic
Neat with Geosynthetic

1200
1150
1100
1050
1000
1/3

1/5

1/3,2/3

Position of Geosynthetic in Relation to Specimen Height

Figure 3.4 Effects of location and stabilization on thegeosynthetics-BCS particle interaction as a measure of elastic
stiffness.
It can be noted that magnitude of the elastic stiffness is
influenced by all the three factors.
4.4 Typical Failure Characteristics
All specimens tested in this Study were loaded to post-failure
conditions and the progressive deformation and failure
characteristics were carefully monitored, observed andinvestigated.
Cored specimens were also adopted for purposes ofanalyzing the progressive deformation and failure modes incomparative terms.

5 CONCLUSIONS
Laboratory investigation was undertaken to study thecharacteristics of the interaction of geomat and fine grained
black cotton soils. In general it was observed that the degree
and extent of the geomat-BCS particle interaction isdependent on several factors mainly including particle size,
texture, shape, moisture-suction variation and location ofembedded geomat relative to the layer thickness defining theactive zone, confining pressures and tensile forces.
The following specific conclusions can be made from theresults reported in this study.
1. The geotechnical engineering properties of black
cottonsoils
arehighlysusceptible
tomoisture-suction variations.
2. The interaction between geomat and black cotton soil
is significantly influenced by the location ofimbedding, mode and type of geosynthetic
reinforcement.
3. Whereas geomats were found to be effective inenhancing the strength and deformation propertiesof fine grained black cotton soils, the influence ofgeogrids was minimal in this case.
4. Geosynthetic reinforcement contributes largely tomobilization and distribution of tensile stresses andcontainment of tensile straining, whilst OPMC
stabilization contributes immensely to theenhancement of compressive stresses and control ofcompressive straining.
5. Where OPMC stabilization contributes significantly
to the enhancement of the magnitude of angle ofinternal friction due mainly to particle
agglomeration,
the
role
of
geosynthetic
reinforcement may appear minimal in the short-term.
However, due to scale and short-term loading effects,
it is important that full scale experimental testing be
undertaken under long-term dynamic loading
simulating various environmental changes andconditions.
6 ACKNOWLEDGMENT
The authors wish to acknowledge the Japan International
Cooperation Agency (JICA), White Nile Oil Exploration

Company, England, and Kajima Corporation, Japan, forsponsoring most of this research as well as the University of
Nairobi, Materials Testing and Research Department,
Ministry of Roads, Kenya and Tensar International, Englandfor their technical support. The authors are also indebted to
Brian Reddings and Sylvester Kotheki for their efforts incompiling the data,
as well as members of the Kenya
Geotechnical Society Research Team for their contributions
during the entire study. The acknowledgement would beincomplete without recognizing the input of Prof. Bernard
Njoroge of the University of Nairobi.
7 REFERENCES[1]

Enni MP, Tatsuoka F, Bathurst RJ, Stevenson PE, Zornberg JG,“Advances in geosynthetics materials and applications for soil
reinforcement and environmental protection works”, EJGE, Bouquet,2008.
[2] Tatsuoka F, Koseki J, Tateyama M, “Performance of earth
reinforcement structures during the Great Hanshin Earthquake”,
Special Lecture, Procs. Int. Symp. On Earth Reinforcement, IS
Kyushu ’96, Balkema, vol. 2, pp 973-1008.
[3] Tatsuoka F, Koseki J, Tateyama M, Munaf Y, Horri N, “ Seismic
stability against high seismic loads of geosynthetic-reinforced soil
retaining structures”, Keynote Lecture, 6th Int. Conf. on Geosynthetics,
1998, Atlanta, vol. 1, pp. 103-142.
[4] Zornberg JG, “Discrete framework for limit equilibrium analysis offibre reinforced soil”, geotechnique, 2002, vol. 52, No. 8, pp. 593-604.
[5] Mukabi JN, “The role of enhanced research in geotechnical engineeringfor pragmatic infrastructure development within Vision 2030”,
Keynote Lecture, Procs. Int. Conf., IEK, 2007, pp. 11-62.
[6] Mukabi JN, “Case Study Analysis of OPMC improved foundation
ground, pavements and geo-structures employing the GECPRO model”,
to be published in Procs. Of ISSMGE International Symposium on
Ground Improvement (IS-GI), Brussels, 2012.
[7] Gichaga FJ, Visweswaranya TG, Sahu BK, “Prediction of swell of
black cotton soils in Nairobi”, Procs. Int. Symp. On prediction andperformance in geotechnical engineering, 1987, Alberta, pp.259-266.
[8] Gono, K., Mukabi, J.N., Koishikawa, K., Hatekayama, R., Feleke G.,
Demoze W., Zelalem A., “Characterization of some engineering
aspects of Black Cotton Soils as pavement foundation materials”, Procsof the Int. Civil Engineering Conference on Sustainable development in
the 21st Century, Nairobi, 2003.
[9] Mukabi JN, Kotheki S, Ngigi A, Gono K, Njoroge BN, Murunga PA,
Sidai V, “Characterization of black cotton soil under static anddynamic loading- Testing and Analyses”, Proc. Int. Conf. on GE, 2010,
Tunis, pp. 67-77.
[10] Mukabi JN, Kotheki S, Ngigi A, Gono K, Njoroge BN, Murunga PA,
Sidai V, “Characterization of black cotton soil under static anddynamic loading- Discussions and Applications”, Proc. Int. Conf. on
GE, 2010, Tunis, pp. 78-89.
[11] Mukabi, J.N., Murunga P.A, Wambura J.H. & Maina J.N., “ Behaviourof con-Aid treated fine grained Kenyan soils”. Geotechnics forDeveloping Africa, Wardle, Blight & Fourie (eds) Balkema, Rotterdam,
1999, ISBN 90 809 082 5.pp.583~519.
[12] Mukabi JN, “Characterization and modeling of various aspects ofpre-failure deformation of clayey Geomaterials – Fundamental theories
and analyses”, Proc. 1st Int. Conf. on GEOMAT, Mie, 2011.
[13] Wekesa S, Mukabi JN, Sidai V, Kotheki S, Okado J, Ogallo J, Amoyo
G, Ngigi L, “Quantitative analysis to verify the theory of soil particle
agglomeration and its’ influence on strength and deformation resistanceof geomaterials”, Procs. 15th ARC on SMGE, Maputo, 2011, pp.
330-335.
[14] Mukabi JN, “Characterization and modeling of various aspects ofpre-failure deformation of clayey Geomaterials – Applications inmodeling”, Proc. 1st Int. Conf. on GEOMAT, Mie, 2011.
[15] Mukabi JN, “Innovative laboratory and in-situ methods of testing ingeotechnical engineering, to be published in the Proc. of the Int. Conf.
of the Inst. of Engineers Kenya , Nairobi, 2013.
[16] Kogi SK, Mukabi JN, Ndeda M, Wekesa S, “Analysis of enhanced
strength and deformation resistance of some tropical geomaterials

through application of in-situ based Stabilization techniques”, to be
published, Proc. 1st Int. Conf. on GEOMAT, Mie, 2011.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Soil Liquefaction Susceptibility Zoning due to Earthquake Using GIS andGeotechnical Data
Bhuyan Md. Habibullah. Rama Mohan Pokhrel. Jiro Kuwano.
Shinya Tachibana. , Department of civil and environmental engineering, Saitama University, Japan.


ABSTRACT: Liquefaction is an earthquake ground failure
mechanism that occurs in loose, saturated granular sediments(particularly sand and silty sand) and has caused extensive damage
to the ground. Liquefaction potential zoning is the process of
estimating the response of soil layers under earthquake excitations
and the variation of earthquake characteristics on the ground surface.
Ground conditions play important roles in the prediction of hazards
caused by earthquake. Thus to evaluate seismic hazards for a widearea, ground formation history along with soil properties must be
known. This paper describes the ground conditions and behavior of
Satte city as a result of Earthquake. In this paper, Geographical
Information System (GIS) is used to obtain soil liquefaction hazard
map. Spatial variations of soil properties are estimated from theavailable borehole locations where SPT N-values, water table depth
and grain size distribution are known. Geomorphological land
classification components such as mountain and hill, terrace, valley
plain, alluvial fan, natural levee, delta, sand dune, reclaimed land,drained land, former river channel etc. are investigated to attain soilliquefaction hazard map. Methodologies of hazard assessment andthe resulting maps will be presented. These maps can be useful forassessing the approximate areas affected by hazards and for disasterprevention planning.

Keywords: Liquefaction potential, PGA, GIS, Satte city.
1. INTRODUCTION
Soil liquefaction is one of the most devastating types ofgeological effects induced by earthquake. It is wellrecognized that numerous engineering structures have been
severely damaged due to liquefaction of the supporting soils
during earthquake.
Japan is situated in a seismically very active zone because of
its location in marginal areas of the Pacific, Philippine Sea,
North American, and Eurasian plates. This is why it is struck
by frequent earthquakes. Past Earthquake history in Japan
resembles it.
These frequent earthquake ultimately affect not only humanbeing but also social, cultural and mostly financial status of
the region. The 1964 Niigata earthquake and 1995 Kobe
earthquake were the most destructive liquefaction effects
observed in Japan [1]. However, at most recent March
11’2011 Gigantic Tohoku Pacific Earthquake was thestrongest to hit Japan and one of the top five largest
earthquakes in the world since seismological record-keeping
began. It was followed by a tsunami with waves of up to 10 m.
The disaster left thousands dead and inflicted extensive
material damage to buildings and infrastructure that led tosignificant accidents at four major nuclear power stations and4,460 people missing.

The Satte city of Saitama Prefecture is targeted as a case study
in this paper. Saitama Prefecture has been affected by several
destroying earthquakes of magnitudes greater than eight in
the past times. The great Kanto Earthquake of magnitude 8.3
in 1923 hit this area which was the latest great one before
March 11’2011 Gigantic Tohoku Pacific Earthquake. The
great Kanto Earthquake devastated Tokyo, the port city of
Yokohama, Surrounding prefectures of Saitama, Chiba,
Kanagawa, and Shizuoka, and caused widespread damage
throughout the Kanto region.
Among the various geomorphological units, the reclaimed
land, drained land, river channel, lowland between sanddunes or bars, marginal part of sand dune, natural levee andbanking area in swampy lowland are most susceptible toliquefaction [2]. From the Satte city geological map in Fig.1,
it is observed that this city is built up in lowland area. Yellow
color shows natural levee, an elongate embankment
compounded of sand and silt and deposited along both banks
of a river channel comprises some major area of the city.
Moreover, river plain and old river plain consist of almost therest of the city.
Around 50 borehole points of Satte city are considered forthis research work. Borehole data analyses are done in Excel
for different PGA and Liquefaction potential Index refer to(6) is found from Excel calculation at final stage. These
values are given as input in GIS according to corresponding
latitude and longitude of borehole points and liquefaction
potential map are originated accordingly. Finally, theobtained liquefaction potential map are evaluated with recent
March 11’2011 Gigantic Tohoku Pacific Earthquakeliquefaction history.
1.1 Study area
According to April 1, 2011, the Satte city has an estimatedpopulation of 54,444, with household number is 21,449 andthe density of 1,603.658 persons per km². The total area is33.95 km². The city was founded on October 1, 1986. The
city is located at 36°04′30″N and 139°43′45″E. Average
altitude of the city is 9 meter. The city is almost downstreamof Tone River. It is at Saitama prefecture in Kanto region and
43 km from Tokyo at N-W corner.
At March 11’2011 Gigantic Tohoku Pacific Earthquake, thehighest PGA observed 3.0 g (USGS 2011) [3] in Japan.
However, at Satte City of Saitama Prefecture, the observed
PGA was 0.3 g [4]. In this paper, the liquefaction potential
has been calculated by using 0.15 g and 0.3 g PGA for Sattecity.

2 METHODOLOGY
Liquefaction potential for Sattecity is evaluated by using thegeotechnical data as type of soil,
SPT N-value, depth of water table,mean grain size and unit weight ofsoil. These geotechnical data werecollected
fromsubsoilinvestigation reports (boreholelog) as shown in Fig. 3. The
borehole data in the study area were collected from different
depths ranging from 10 m to 50 m below ground surface.
However, in this study, borehole data parameters are
considered up to 20m from ground surface. It is consideredthat no liquefaction affect on ground for soil properties below
20m depth.
According to [5], an ability to resist the liquefaction of a soil
element at an arbitrary depth may be expressed by theliquefaction resistance factor (FL) identified by the following
equationR(1)
L
When the factor FL at a certain soil is less than 1.0, weFL evaluated that the soil liquefies during earthquake.
In equation (1), R is the in situ resistance or undrained cyclic
strength of a soil element to dynamic loads during
earthquakes. The value of R is determined by the following
equations based on mean particle diameter (D50) as follows:for 0.04mm  D50  0.6mm

R  0.0882

N
0.35
 0.225 log10
D50
v`0.7

for 0.6 mm  D50  1.5 mm

Fig.1 Satte city Geological Map

R = 0.0882

N 0.05v`0.7(3)

Where N is the number of blows of the standard penetration
test, v` is the effective overburden pressure (in kgf/cm2) andD50 is the mean particle diameter (in mm) illustrated in Table
1.
Due to seismic motion, the dynamic load induced in the soil
element can be estimated by the following equation:

max  s max v

rd
(4)
g v`
v`
Where max is the maximum shear stress (in kgf/cm2),L s max is the maximum acceleration at the ground surface (ingals), g is the acceleration due to gravity (=980 gals), v isthe total overburden pressure (in kgf/cm2), and rd is the
reduction factor of dynamic shear stress to account for the
deformation of the ground. Iwasaki et al. proposed thefollowing equation of rd from a number of seismic response
analyses of the ground:(2)

rd = 1.0  0.015Z(5)

Where Z is the depth in meters.
It is evident that the damage to foundations of structure due tosoil liquefaction is considerably affected by the severity ofliquefaction. An index of liquefaction potential, PL , can beintroduced to express the severity of liquefaction as follows:20

PL   F  W ( Z )dZ(6)0

Where F = 1.0  FL for FL  1.0, F= 0 for FL > 1.0 and
W (z) =10  0.5Z (Z in meters) demonstrated in Fig. (2). For
the case of FL = 0.0 for the entire range from Z= 0 to
20meters, PL becomes 100, and for the case of FL  1.0 for
the entire range from Z= 0 to 20meters, PL becomes 0.
Liquefaction potential calculation Excel sheet sample isattached herewith in Table 2. Stress at every 1m is calculated
based upon soil type and average density and the following
overburden stress is calculated. Based upon the depth ofwater level from ground, pore water pressure is calculated
and effective stress is found subtracting pore water pressure
from overburden pressure. Subsequent that, reduction factor
for dynamic shear stress rd is considered based on depth
according to (5). Afterward the in situ resistance of soil R is
considered conferring (2) or (3) based on mean particle
diameter attached in Table 1. In next step, dynamic load
induced in the soil element by seismic motion, L is intended
according to (4). Finally an ability to resist liquefaction
FL and Liquefaction potential index PL is calculated
according to (1) and (6) respectively.
Liquefaction potential Index, PL values of borehole data areplotted in GIS for 150 gal and Fig. 4 is the representation of
that. This zoning map shows that most of the area has low risk
for liquefaction according to guideline proposed by [5]. The
rest 30% of the borehole points show high risk ofliquefaction.

Fig. 2 Definition of PL
Table 1 Average values of the Unit Weights and Meanparticle Diameters of Different Type of Soil.(This table wasused only when these values were not tested) [5]Soil TypeSurface soil
Silt
Sandy Silt
Silty Sand
Very fine Sand
Fine Sand
Medium Sand
Coarse Sand
GravelUnit(t/m3)
1.7
1.75
1.8
1.8
1.85
1.95
2.0
2.0
2.1

WeightMean Particle
Diameter, D50 (mm)
0.02
0.025
0.04
0.07
0.1
0.15
0.35
0.6
2.0

Table 2: Calculation of Liguefaction Potential Value for one borehole of PGA 300 gal
Depth(m) Soil Type
1.3
2.3
3.3
4.3
5.3
6.3
7.3
8.3
9.3
10.3
11.3
12.3
13.3
14.3
15.3
16.3
17.3
18.3
19.3
20

Avg. Density g Stress,1m Ovd. Pres. D fr. WT Pore Pres.Eff. stress kgf/cm^2 N

surface soil
siltsand and silt
silt
silt
clayey silt
clayey silt
clayey silt
clayey silt
sandy silt
sandy silt
sandy silt
sandy silt
sandy silt
sandy silt
silt
silt
silt
silt
silt

1.7
1.7
1.7
1.7
1.6
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7
1.7

9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8
9.8

16.66
16.66
16.66
16.66
15.68
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66
16.66

21.65838.31854.97871.63887.318103.978120.638
137.298
153.958
170.618
187.278203.938220.598237.258
253.918
270.578287.238303.898
320.558
337.218

0
0.15
1.15
2.15
3.15
4.15
5.15
6.15
7.15
8.15
9.15
10.15
11.15
12.15
13.15
14.15
15.15
16.15
17.15
17.85

0
1.47
11.27
21.07
30.87
40.67
50.47
60.27
70.07
79.87
89.67
99.47
109.27119.07128.87138.67148.47158.27168.07174.9321.65836.84843.70850.56856.44863.30870.16877.02883.88890.74897.608104.468111.328118.188
125.048131.908
138.768
145.628
152.488
162.288

0.21879 4
0.37224 1.7
0.44154 5
0.51084 00.57024 00.63954 00.70884 0
0.77814 00.84744 00.91674 00.98604 0
1.05534 0
1.12464 0
1.19394 0
1.26323 0
1.33253 0
1.40183 0
1.47113 01.54043 0
1.63943 0

rd
0.981
0.966
0.951
0.936
0.921
0.906
0.891
0.876
0.861
0.846
0.831
0.816
0.801
0.786
0.771
0.756
0.741
0.726
0.711
0.7

R

PGA

0.4637 300
0.368 300
0.4425 300
0.2579 300
0.2579 300
0.2579 300
0.2579 300
0.2579 300
0.2579 300
0.212 300
0.212 300
0.212 300
0.212 300
0.212 300
0.212 300
0.2579 300
0.2579 300
0.2579 300
0.2579 300
0.2579 300

L

FL

IL

0.30020.30740.366
0.40570.43590.45530.46870.4777
0.48340.4866
0.48780.48730.4856
0.48270.47890.47440.4692
0.46350.45720.4453

1.54491.1972
1.20890.6356
0.5916
0.56640.55020.53980.53340.43560.43450.43490.43650.43910.44250.5436
0.5496
0.55640.564
0.57920002.86033.00162.96992.85612.69212.49622.73762.45992.1756
1.88771.59861.31
0.84440.608
0.377
0.1526031

The following guideline proposed by [5] is used to assess theliquefaction potential.

PL = 00< PL  5
5< PL  15
15< PL: Liquefaction risk is very low: Liquefaction risk is low: Liquefaction risk is high: Liquefaction risk is very highFig. 3 Typical borehole log of Satte city

Fig. 4 Liquefaction Potential at borehole locations of Satte city at PGA 150 gal

Fig. 5 Distribution of peak ground acceleration provided by
NIED, ERI (The University of Tokyo), AIST, and PARI

uniform, and contained several asperities radiating strong
ground motions. Another point to note is that Tokyo and its
surrounding area near the bottom of the Fig. 5 were subjected
to strong shaking. Therefore, damage occurred at many
places therein. Moreover, according to latitude and longitudeof Satte city (36°04′30″N and 139°43′45″E), it is observedthat Satte city of Saitama Prefecture faces 300 gal of PGA(Fig. 5) during the recent March 11’2011 Gigantic Tohoku
Pacific Earthquake.
According to Fig. 5, the observed PGA (300 gal) for Satte city
is applied in GIS to observe liquefaction and it is observedthat all the borehole points show high to very liquefaction
values ranges from 7 to 51 as shown in Fig. 6. Satte and Kuki
city sites are visited just after the March 11’2011 Gigantic
Tohoku Pacific Earthquake. Sand boiling through the groundand paved road, house and electric post tilting, manholes
coming up the ground are the consequences of liquefaction
which match with the analysis attached in Fig. 6. Moreover,
Seismic Damage and Liquefaction history Map at March
11’2011 Tohoku Pacific Earthquake attached in Fig. 7
evaluates high to very high risk of liquefaction.

The distribution of peak ground acceleration (PGA) in Fig. 5
implies that the rupture process during the earthquake was not

Fig. 6 Liquefaction Potential at borehole locations of Satte city at PGA 300 gal

Fig. 7 Seismic Damage and Liquefaction history map at March 11’2011 Tohoku Pacific Earthquake [3]3 CONCLUSION
Geotechnical problems are made up of liquefaction in
foundations and associated ground deformations. Due todamage effect, the post-earthquake response and restoration
works have been made difficult and delayed. In this paper,around 50 borehole data are analyzed to draw liquefaction
potential map for Satte city. In case of 150 Gal PGA, theborehole points display low to medium risk for liquefaction
whereas for 300 Gal PGA, the borehole points display high tovery high risk for liquefaction which is a representation ofrecent 2011 Tohoku Pacific Earthquake. These features havenot been considered in the conventional design philosophy
and should be discussed from now on. Geotechnical Damage
due to the 2011 Christchurch, New Zealand Earthquake and2011 Tohoku Pacific Earthquake in Japan prompts us
considering this true fact. People are now aware of groundsurface and this is why geotechnical engineers need to work
hard now a days. These maps can be useful for assessing theapproximate areas affected by hazards and for disasterprevention planning.

4 ACKNOWLEDGMENT
The first author acknowledges to the Asian Development
Bank (ADB)-JSP for its support by providing scholarship forthis research work.
5 REFERENCES
[1]
[2]

[3]
[4]

[5]

Pokhrel, R.M., Kuwano, J., Tachibana,S., “Liquefaction hazard
zonation mapping of the Saitama City, Japan,” Journal of Nepal
Geotechnical Society, vol. 40, 2010, pp. 69-76
Kotoda, K., Wakamatsu, K., and Oya, M., “Mapping Liquefaction
Potential based on Geomorphological land classification,” inProceedings of Ninth World Conference on Earthquake Engineering,
August 2-9,1988, Tokyo-Kyoto, JAPAN(Vol. III).
http://earthquake.usgs.gov/earthquakes/eqinthenews/2011/usc0001xgp
Towhata et al., “Earthquake News on Gigantic Tohoku Pacific
Earthquake in Japan,”April’2011, ISSMGE Bulletin, Volume 5, 2011,
Issue2, page 46-66.
Iwasaki, T., Arakawa, T., Todika, K., “Simplified Procedure forassessing Soil Liquefaction during Earthquakes,” in Proc. International
Conference on Soil Dynamics and Earthquake Engineering, 1981

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Evaluation on the Results of Multistage Shear Test
Hormdee D., Kaikeerati N., Department of Civil Engineering, Khon Kaen University, Thailandcell pressure. The total strain that the sample was subjected toduring MCU test amounted to about 25%.

ABSTRACT: Shear strength is a very important soil property todetermine the stability of foundations, retaining walls, slopes andembankments. In order to solve those problems in geotechnicalengineering, shear strengths are usually determined from laboratory
tests performed on specimens prepared by compaction in thelaboratory or undisturbed samples obtained from exploratory soil
borings. In order to determine the shear strength parameters, the soil
specimens are required at least 3 identical samples. To eliminate theeffects of soil variability on the result, multistage testing is used tomaximize the amount of shear strength information that can beobtained from only one specimen by using several confiningpressures. Test results demonstrate that is possible and convenient toperform multistage shear test on compacted soil to measure shear
strength. The direct shear tests were carried out using the multistage
technique and the results are quite well comparable to those of
traditional shear tests.

To avoid subjecting the sample of high strain, Sridharan andNarasimha Rao [3] investigated a new approach to multistage
testing, which made use of Kondner's [4] hyperbolic
stress-strain relationship for predicting the state of stress at
failure from the stress-strain and pore-water pressure-strain
curves of soils can be approximated by rectangular hyperbola,
for which the equation is shown in Eq.1.
/(1)

where:
shear strain
shear stressy-interception of / and relationship
1/ (from asymptotic value)slope of / and relationship
1/ (from asymptotic value)
initial tangent modulus andfailure shear strength, with the normal stress,

Keywords: Multistage Test, Shear Strength Parameters,
Hyperbolic Stress-Strain Relationship, Direct Shear Test
1. INTRODUCTION
The most widely used laboratory equipments forinvestigating the strength and deformation behavior of soils iseither the triaxial or direct shear apparatus. The apparatus isversatile and can be used for the measurement of many
parameters, including shear strength characteristics,
consolidation characteristics and the permeability of soil. Its
key features include facilities for the control of the magnitude(but not direction) of the principal stresses, the control ofdrainage, and the measurement of pore pressures [1].

Similarly for predicting pore-water pressure ( ), Sridharanand Narasimha Rao [3] used Eq.2.
/(2)

where:
y-interception of / and relationshipslope of / and relationship
1/ . andpore-water pressure at failure

In traditional tests, each specimen undergoes a phase of
consolidation and shearing and thus supplies a single stressversus strain trend and of course only one state of stress at
failure. A series of three or more specimens, consolidated atvarious stress levels, supplies an ensemble of stress data
allowing identification of a failure envelope and thus shear
strength parameters. Sometimes it is impossible to have a setof homogeneous or identical re-compacted specimens, foreconomic reasons or because some soils formations may bedifficult to sample. In such circumstances the multistage
shear testing technique appears to be an attractive alternative.
This technique has been successfully used to determine shear
strength parameters by just one sample.

The multistage test suggested by Sridharan and Narasimha
Rao [3] referred to as NMCU tests consisted of: (1) assessing
whether the soil lends itself to extrapolation using Kondner’s
method, (2) consolidating the sample to a cell pressure andshearing it to 2 to 4% axial strain, (3) repeating theconsolidation-shear sequence at two higher cell pressure, (4)using Kondner’s method to predict failure stressed at eachcell pressure, and (5) using the results for all three cellpressure together to determine shear strength parameters.
Sridharan and Narasimha Rao [3] report success using thismethod, however using Kondner’s method the value ofpredicted were higher than actual. Sridharan andandNarasimha Rao [5] had suggested that in extrapolating datausing Kondner’s method better prediction is possible if one
extrapolates only to a finite value of strain using (1) and (2)rather than to the asymptotic value of infinite strain. They had
tentatively suggested using a strain of 15%

Kenney and Watson [2] conducted multistage undrained tests(MCU). They report reasonable consistent between c' and '
obtained from MCU tests with the values obtained by using
conventional undrained tests (CCU). MCU tests wereconducted by consolidating and loading it until about failure
strain, then repeating the process twice again with two higher

93

In this paper, the series results of single stage testing werecarried out in direct shear apparatus to compare with theresults of multistage series. Single stage tests were conducted
till 8-15% strain on four samples at confining pressures of100, 200, 300 and 400 kPa. And the loading rate was 0.002%
strain per minute for direct shear test as drained condition
required.

Even though the multistage testing technique has been usedwith some success in the past to determine shear strengthparameters, to define undrained shear strength [6],[7] oreffective shear strength both in unsaturated [8],[9] andsaturated soils [3],[10] in triaxial apparatus and in direct shear
apparatus [11] but still lacking in concerning and carrying out
on the drain condition of testing especially in direct shear
apparatus.

2.3 Multistage Procedure
A multistage test induces more than one consolidation andshearing on the same soil specimen loaded inside a direct
shear box. Specimen preparation and initial test stages ofsaturation are identical to those of traditional direct shear.
Especially the saturation stage in triaxial test, it is required
more or less 5 days to gain the value of B (pore pressure)parameter more than 0.9 and to finish the settlement in thisprocess. This method enables substantial homogeneity of theresults and appreciable time consuming.

In this paper, a series of single and multistage testing is
performed using direct shear apparatus as drained condition
to investigate on the shear strength parameters of the single
stage-multistage comparative study. The preparation of thesoil specimens is controlled as the field condition by using
field dry density and optimum moisture content (OMC).
2. METHODOLOGY
2.1 Preparation of Specimens
The red Khon Kaen loess soil sample was collected as
disturbed sample at a depth of 3-4 m. The index properties
and compaction characteristics are shown in Table.1.

After saturated and consolidated specimen in stage I isfinished, in the shearing stage the specimen is sustained to
produce a significant amount of shear stress. Then thespecimen is released from horizontal stress. To start the nextstage, the specimen is subjected to a higher normal stressbefore the following shearing stage takes place.

Table.1 Properties of soil___________________________________________________
Property
Khon Kaen loess soil___________________________________________________In this paper, the strain for each compression stage has been
carried up to only 3% in each test. In all cases, the peak valueof shear strength are expected for strain values greater than
8%, as is the case of many clayey and silty-clayey soils of
medium to low plasticity [7] and also as is the case of many
not dense condition of compacted soil. The normal stress andloading rate in multistage test were used the same as a series
in single stage test. In the results of direct shear test, the test
was conducted to not failure but carried to 3% of strain ineach stage, and the ultimate peak values are predicted bymaking use of Kondner's hypothesis [4].

Specific gravity (Gs)
2.64
Liquid limit (LL)
21.2%
Plastic limit (PL)
14.3%
Unified Soil Classification System (USCS) SM-SC
OMC
9.25 %
d max2.02 g/cm3
d,field
1.81 g/cm3___________________________________________________The experimental program was a parametrical study aiming atstudying the effect of important reconstituted conditions ofsoil such as initial water content, initial dry density on shear
strength parameters. As using disturbed samples, they wereprepared in ‘identical’ fashion by controlling initial watercontent as OMC and initial dry density as the field dry density(that is about 89.6% of compaction) by using hydraulic jackslowly pressing to 60 mm in squared width and 20 mm inheight.

3 RESULTS AND DISCUSSION
Direct shear tests were carried out in both single stage andmultistage test saturated specimens with normal stresses of100, 200 ,300 and 400 kPa. The single direct shear testing
result on saturated condition is shown in Fig.1.
In consideration of the hyperbolic stress-strain relationship,
the hyperbolic model represents the nonlinear stress-strain
curve of soil using hyperbola. It can be seen that transforming
the hyperbolic equation result in a linear relationship between/ and, shown in (1). The validity of Kondner's
hypothesis is shown in Fig. 2. The stress dependentstress-strain behavior of soil is represented by varying theinitial tangent modulus, , and the failure shear strength,with the normal stress, .

2.2 Single Stage Procedure
To perform the shear strength testing, a two-stage loading
procedure was used in each of these tests. In the first stage, aconsolidation was applied. In the direct shear test, theconsolidation was applied by applying a vertical load to thehorizontal plane, becoming the eventual failure plane. In
traditional tests, each specimen undergone a phase of
consolidation and shearing and thus supplied a single stressversus strain trend and of course only one state of stress at
failure [7]. A series of 3 or 4 specimens, consolidated atvarious confining stress levels, supplied an ensemble of stressdata allowing identification of a failure envelope and thus
shear strength parameters.

These plots do produce straight line relationships and as such
indicate that the behavior of soils lends itself to prediction by
Kondner's technique. The failure shear stress ( ) for single

94Since Kondner's technique was found to be valid for Khon
Kaen Loess soil, it is possible to conduct the multistage direct
shear test to predict failure stresses. Fig.3 presents the plot of/ versus for multistage direct shear test carried up to
3% strain in each stage that are predictably linear Table.3
presents the values of predicted from asymptotic value. As
can be seen from the Fig.4, the predicted results of strengthparameters in multistage test are relatively validity by giving
the quite lower value of internal friction angle (about 4% oferror) with no cohesion. So the prediction of failure stresses
using Kondner's relation is nearly perfect when results are
extrapolated to a value of finite strain determined byanalyzing data of single stage test.

tests have been predicted from asymptotic values and using(1) for failure strains of 10, 15, and 20%.
400

100 kPa
200 kPa
300 kPa
400 kPa

300
200
100
0
0

2

4

6

8

10

0.08

0.06

s/

0

-2

n(kPa)
100
200
300
400

a value
0.01400.00830.00700.0028b value0.01590.00650.00380.0032100 kPa

0.04
200 kPa-4

0.02

300 kPa
400 kPa

-6

0
0

1

2

Fig.1 Testing results in single stage direct shear test
0.16

0.12

n(kPa)
100
200
300
400

a value
0.03290.01740.00410.0090b value0.00830.00420.00350.002334s (%)
Fig.3 Hyperbolic representation of stress-strain curve in
multistage direct shear test

100 kPa

500

0.08

Shear strength, kPa

s /

Single Stage Test
200 kPa
300 kPa
0.04
400 kPa
0.00
0

5

10

400

Multistage Test

300
c' = 17.46 kPa
' = 38.2o
200c' = 0 kPa
' = 36.6o

100

15

s (%)

00

Fig.2 Hyperbolic representation of stress-strain curve insingle stage direct shear test

100

200

300

400

500

Normal stress, kPa

Fig.4 Failure Envelope in Direct Shear Test

The predicted and experimentally obtained values of arepresented in Table. 2. Also presented in this table is value ofpercentage agreement, which is the ratio if predicted toexperimentally obtained values. A study of Table. 2 showsthat the value of predicted by using approaching infinity
is always higher than experimental values. Agreement withexperimental value is better when value of strain used is about
15%.

4

DISCUSSION AND CONCLUSIONS

Testing results demonstrate that is possible and convenient toperform multistage shear test on compacted soil to measure
shear strength. The drained direct shear test was carried out
by using the multistage technique and, as shown, the results
are quite well comparable to those of traditional shear tests.
The proposed multistage shear testing procedure can be used95

to evaluate the shear strength parameters from a single test.
Table.3: Predicted Values of f for Khon Kaen Loess Soil
Experimental Value of f Obtained
n,Value of at 3% from Asymptotic
kPa
Strain, kPa
Value, kPa

in single stage direct shear tests.

100
200
300
400

5

50.4
115.3
166.1
265.2

62.9
153.8
263.2
312.5

Predicted Value of f
Using (1) for Shear
Strain of 15%, kPa
59.4
141.8
234.4
297.0

ACKNOWLEDGEMENT

The authors would like to greatly acknowledge the Thailand
Research Funding for their sponsorship. The help rendered bysenior project’s students and other staffs at Department ofCivil Engineering, Khon Kaen University in preparing thesoil samples is acknowledged.
6 REFERENCES
Head, KM, Manual of Soil Laboratory Testing. Vol.3: Effective Stress
Tests. London, UK Pentach Press, 1986.
[2] Kenny, TC and Watson, GH. "Multi-stage Triaxial Test for
Determining c' and f' of Saturated Soils," in Proceedings of the 5thInternational Conference on Soil Mechanics and Foundation
Engineering, Vol. 1, Praris, France, 1961, pp. 191-195.
[3] Sridharan, A and Narasimha Rao, S, "A New Approach to Multi-stage
Triaxial Test," Journal of Soil Mechanics and Foundation Division,
Proceedings of the American Society of Civil Engineer, Vol. 98, No.
SM11, Nov. 1972, pp. 1279-1286.
[4] Kondner, RL, "Hyperbolic Stress-Strain Response, Cohesive Soils,"
Journal of Soil Mechanics and Foundation Division, Proceedings of
the American Society of Civil Engineer, Vol. 89, No. SM11, Jan.
1963,pp. 115-143.
[5] Sridharan, A and Narasimha Rao, S, "Hyperbolic Representation ofStrenght, Pore Pressure and Volume Changes with Axial Strain in
Triaxial Tests," in Proceedings of the Symposium on Strength andDeformation Behavior of Soils, Vol. 1, Bangalore, India, 1972, pp.
33-42.
[6] Anderson, WF, “The Use of Multi-Stage Triaxial Tests to Find theundrained Strength Parameters of Stony Boulder Clay,” in Proceedings,
Institution of Civil Engineers, Pt. 2, Vol, 57, Thomas Telford Ltd.,
London, 1974, pp. 367-373.
[7] Soranzo, M, “Results and Interpretation of Multistage Triaxial
Compression Test”, Advanced Triaxial Testing of Soil and Rock,ASTM STP 977, ASTM, Philadelphia, 1988, pp. 353-362.
[8] De Beer, EE, “The Cell Test”, Geotechnique, Vol.2, No. 2, 1950, pp.
162-172.
[9] Ho, DYF and D. G. Fredlund, DG, “A Multi-Stage Triaxial Test for
Unsaturated Soils”, Geotechnical Testing Journal, Vol. 5, No. 12, 1982,
pp. 18-25.
[10] Kenny, TC and Watson, GH “Multiple-Stage Triaxial Tests for
Determining c’ and ’ of Saturated Soils”, in Proceedings, Fifth
International Conference on Soil Mechanics and Foundation
Engineering, Paris, Vol. 1, 1961, pp. 191-195.
[11] Gan, KY and Fredlund, DG, “Multistage Direct Shear Testing of
Unsaturated Soils”, Geotechnical Testing Journal, Vol. 11, No. 2, 1988,
pp. 132-138.

Table. 2: Predicted and Experimental values of

[1]

96

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Verification of Effects of Countermeasures for Seepage of River Levees by
On-site Monitoring of the Water Level of Levees


Hiroyuki MASUYAMA1. Yukiko SAITO2. Masanori ISHIHARA3. Tetsuya SASAKI4. Hirotoshi MORI5
1,2,3,4, Soil Mechanics and Dynamics Research Team, PWRI (Public Works Research Institute)
5, Policy Bureau, Ministry of Land, Infrastructure, Transport and Tourism

Some interesting results are introduced in this study.
ABSTRACT: Japan is a country where many floods occur.
Therefore, importance is attached to safety with respect to seepage inJapan. There are however, a few cases where the performance ofcountermeasures has been confirmed at rivers. To verify the effectsof countermeasure methods with respect to seepages at rivers,
on-site monitoring and three-dimensional seepage flow analyses
were executed during this study. The monitoring of landsidecountermeasures targeted drains. Drains are typical measurements inlandsides. The monitoring of waterside countermeasures targeted
sheet piles and seepage control sheets.
As a result, the effects of countermeasures were confirmed byon-site monitoring.

2. COUNTERMEASURES TO TARGET
Tab-1 shows the list of targeted countermeasures.
Monitoring of landside measures targeted three different
drains. Monitoring of waterside measures targeted theseepage control sheets and sheet piles. Fig-2 shows thelocation map of the targeted site.
Tab-1 List of countermeasures to target

Keywords: river levee, on-site monitoring, drain, sheet pile,
three-dimensional seepage analysis

No.
Measuresposition

1. INTRODUCTION

Measures
Cross
SectionNo.
Measuresposition

Improvement of safety with respect to seepage is advanced
in order of 1.checks for safety, 2. extraction of weak point
portions, 3.selection of countermeasures, 4. construction,
5.monitoring, and 6.maintenance.
It is necessary to prove certainty with respect to the effectsof countermeasures for river levees so as to advance theirreinforcement. The effects of countermeasures in the crosssection are shown in Fig-1.

1

2

Landside ( Drain A )Landside ( Drain B )34Landside ( Drain C )Waterside ( Sheet pile )Measures
Cross
SectionFig-1 Effect of countermeasures in cross section
It is important to monitor the on-site water levels of riverlevees. The authors have compiled “The Water level of thelevees observation manual” [1], and have observed the waterlevel of river levees on-site. Additionally, Authors haveclarified a three-dimensional complex seepage flow of riverlevees [2]–[5].

Fig-2 Location map of targeted site97

3. OBSERVATION HOLE
An observation hole in the
cross section is shown in
Fig-3.
Observation holes use hole
up dug by boring. A tube is set
up in the hole up dug byboring.
The hydraulic pressure typewater meter is made to hang in
the tube and is set up. The
strainer is installed in thevicinity of the water meter.
Mud accumulation is installed
under the hole.
Fig-3 Observation hole in cross section

Fig-6 Monitor part in the cross section
5.2k+180m)(section of

Observations were carried out as of August, 2009, and arecontinuing into August, 2011. The observation frequency of
the water level of the levee is made for other periods once
every five minutes and once every two minutes during theflood season from June to October. Additionally, water
meters of rivers and rain gauges were set up.
Floods with a large influence on the water level of the levee
have not occurred at the AIRA River since observations at the
beginning of August, 2009. Fig-7 shows the observational
results of June 19th, 2010 where water levels rose most thanin any other period. Water levels of levees rise as a result of
rainfall. The water level of levees in sections with drains islower than the water level of levees in the section without
drains. As a result, the effects of the drain can be confirmed.
Continuous monitoring is scheduled for the future. The
effects of the drain are confirmed.

4. MONITORING OF EFFECT OF LANDSIDEMEASURES (Drain A)
Aira River levees are composed of sandy soil (coefficient of
permeability ks=1.0E-5m/s), including Shirasu, which is apyroclastic flow deposit that is peculiar to the region on thefoundation subgrade in the layer of sand. PWRI executed theinstallation and data observation of the observation hole.
The plan, the geological features profile, and the crosssection are shown in Fig-4, 5 and 6. In the monitor section,
there is a section where the seepage control sheet is used
together with the drain.

Fig-7 Observational results of water level of the levee on
June 19th, 2010

Fig-4 Monitor part in plan

5. THE MONITORING OF EFFECTS OF LANDSIDEMEASURES (Drain B)
The on-site monitoring of drain measures was done in thevicinity of the YODO River water system MO River 0.9k
points. Fig-8 shows the plan of the observation hole.
The monitoring results, as shown in Fig-9 and 10, wereobtained in the small flood that occurred on July 19th in 2006.
The flood that showed the effect of the drain for theobservation period did not occur.

Fig-5 Monitor part in profile (position of levee crown landside shoulder)98

Fig-12 Water level of levee in cross section
Fig-8 observation hole in plan
6. THE MONITORING OF EFFECTS OF LANDSIDEMEASURES (Drain C)
The on-site monitoring of drain measures was carried out inthe vicinity of the YODO River water system at SETA River
68.6k points. Fig-13 shows the plan of the observation hole.
Fig-14 shows the levee in cross section.

Fig-9 Observational results of water level of the levee andon July 19th, 2006

Fig-13 the plan of the observation holeFig-10 Water level of the levee in the cross section at themonitoring site at 8:00 on July 19th, 2006
The seepage flow analysis on two dimensions was done
based on the observational results. Fig-11 shows theanalytical model chart. It also analyzed no measures and theeffects of measures were verified. Fig-12 shows the effects ofmeasures of the drain that were obtained in the analysis.

Fig-14 SETA river levee in cross section

84.5

0

84.0

10

83.5

20

83.0

30

82.5

40

82.0

50

81.5

60

81.0
6/15 0:00

Fig-11 Analytical model chart

No.1

6/16 0:00
No.2

No.3

6/17 0:00
No.4

6/18 0:00

6/19 0:00

Water level of SETA River

Amount of Rainfall per hour (mm/hr)

Water Level(OP.m)

When flooded by discharging the SETA River ARAI Weir,
the rising of the water level of the levee was generated for theobservation period(fig-15).

70
6/20 0:00
RainfallFig-15 Observational results of the water level of thelevee on June 15th - 20th, 2006
99

The seepage flow analysis on two dimensions was carriedout based on observational results. Fig-16 shows theanalytical model chart.

Fig-19 Monitor part in profile(position of levee crown landside shoulder)Fig-16 Analytical model chart
It also analyzed no measures. The effects of measures wereverified. Fig-17 shows the effect of measures of the drain
obtained by the analysis.

Fig-17 Water level of the levee in the cross section

Fig-20 Monitor part in cross section (section of 6.4k-14m)

7. THE MONITORING OF EFFECTS OF WATERSIDEMEASURES (SHEET PILE)
7.1 On-site monitoring
1) Monitoring procedures
The seepage control sheet and sheet pile were set up in thetoe of the levees at the JYOBARU River. It was set it up in the
80m section. The plan, the geological features profile, and therepresentative crossing section are shown in Fig-18-20. The
JYOBARU River levees are composed of decomposed
granite soil (coefficient of permeability ks=7.0E-4m/s of Bs1
and coefficient of permeability ks=5.0E-5m/s of Bs2) thatweathers the Mesozoic granite on the foundation subgrade of
ARIAKE clay that piled up thickly.
Since October, 2009, observation data has been collected so
far for one year or more. This is continued in August, 2009.
The observation frequency of the water level of the levee isonce an hour. Additionally, Public Works Research Instituteset up a water meter for the river and a rain gauge in October,
2010, beginning observations.

2) Monitoring results
Fig-21 shows the time history from 2010/7/11 to 2010/7/15.
The water level exceeded the estimated high-water level at 8
and 9:00AM on the morning of July 14th. After beginningobservations, the maximum water level that had been
observed at the HIDEKI Bridge was high in second during the
middle of this flood. The water level of the river levee wasconfirmed to follow roughly that in Fig-21, the water level ofriver. The water levels of levees of 0tu and 110tu in theoutside measurement section are high. Additionally, the waterlevels of levees of 90tu have risen within the measurement
section.

Fig-21 Observational results of water levels of the levee on
July 11th - 16th, 2010

Fig-18 monitor part in plan
100The water level of the levee at the levee crown landsidepaste shoulder position at the time of 9:00 when the waterlevel of the river reached the peak water level on the 14th isshown in Fig-22. As for the water level of the levee in thesection where the measures with sealing sheet piles and theseepage control sheet were executed, it was lower than thesection without measures by more than 1m. The effects ofmeasures were confirmed with Fig-22. Moreover, the waterlevel has not decreased in the upstream edge in the measuressection. Therefore, it has been understood that there is asection where the effect of measures decreases by athree-dimensional seepage flow.

based on soil profiles (Fig-19 etc.) and soil test results. The
measures reflect the shape of ground as of July, 2010. The
modeled seepage control sheet and sealing sheet pile on thewaterside with the levee crown pavement are shown in Fig-25.
Penetration characteristics of measures were set respectively
as 10cm in thickness based on technological standards [6].
The coefficient of permeability was converted into a value of
10cm in thickness.

Fig-24 Analytical model in the vicinity of river levees
Fig-22 water level of levee at flood peak
7.2 The three-dimensional seepage analysis
The waterside measures followed sheet piles and theseepage control sheet in the vicinity of the CHIKUGO River
water system and JYOBARU River 6.4k points.
1) Analytical model
The range where an analytical model has been built is shownin Fig-23. In the direction of the determination of the lengthand breadth of the levee, the range of an analytical model was
made in the range of 600m×300m. This includes the range inwhich on-site monitoring is carried out. The curve of the river
channel is not seen within the range of the analysis. Therefore,
the shape of an analytical model was made into a rectangle.

Fig-23 Range of modeling of analysis
The soil layer composition and coefficient of permeabilityof the levees and the foundation subgrade were set as shownin Fig-24. This is based on the results of the survey during the
emergency of the 2009/07/26 disaster that were restored

Fig-25 Analytical model in measures section
The rainfall and water level of the river in 2010/7/11 2010/7/15 were set based on observational results. However,the seepage flow reacted in hypersensitivity when the rainfall
condition was given. The analysis was emanated. Therefore,
the analysis was executed without giving the rainfall.
2) Analytical result
The comparison between analytical results andobservational results is shown in Fig-26. This is with respect
to the water level of the levee of the monitored part. Fig-26 isa time history of the water level of levees of 0tu, 90tu, and110tu. Changes in the water levels of the levee for one weekfrom 7/10 show roughly the same tendencies as this Fig withan analytical value and an observed value. Therefore, it isthought that this analytical model and technique roughlycatch the seepage flow.

101

Water level of levee (TPm)

7.5
7.0

8. CONCLUSION

6.5
6.0
5.5
5.0
4.5
4.0
7/10

7/11

7/12

7/13

7/14

7/15

7/16

7/17

0tu observed value（outside measures section)
0tu analytical value(outside measures section)
90tu observed value(inside measures section：vicinity of edge)
90tu analytical value(inside measures section：vicinity of edge)
110tu observed value（outside measures section)
110tu analytical value（outside measures section)

Fig-26 Time history of water level of the levee (outside
measures section and edge)
The profiles of the water levels of the levees at the flood
peak are shown in Fig-27. The water level of the levees in themeasures section lowered relatively more than the water levelof levees outside the measures section. Therefore,
countermeasures for sealing sheet pile and the seepage
control sheet can be confirmed.
Moreover, the flow of the direction of the running water inthe river levee was seen from the flow velocity vector shownin Fig-28 at the edge in the sealing sheet pile section. The
effect of infiltration measures were confirmed as well as themonitoring results. As a result, the influence of
three-dimensional seepage flow at the edge of the measuressection was reproduced during this analysis.

To verify the effect of infiltration measures of the river levee,
on-site monitoring and three-dimensional seepage flow
analyses were executed.
As a result, the effect was confirmed by the observation of
the water level of levees and analytical results with respect tothe drain section of landside measures.
Additionally, the effect of measures for the seepage control
sheet for the water side measures and the sheet pile section
were achieved by the observation of the water level of levees
and the analysis.
Moreover, the situation in which three-dimensional seepage
flow was caused, was confirmed at the edge in the measuressection.
It is important to execute on-site monitoring continuously inthe future so as to confirm the effect of long-term infiltration
measures based on the results, and to make the best use ofsuch for methods of managing design methods and formaintenance.

9. ADDRESS OF THANKS
For cooperation at the OSUMI River national road office, in
the INAGAWA River Office, the BIWAKO River Office,and the CHIKUGO River Office in on-site monitoring.
Rainfall data was offered to the KANOYA City AIRA
Synthesis Branch Office. Thank you very much.

10. REFERENCES
[1]

[2]

Fig-27 Profile of water level of the levees (landside paste
shoulder)

[3]

[4]

[5]

[6]

Fig-28 Flow velocity vector plane on sealing sheet pile
upstream side in the vicinity of GL-1.5m
102

The Cooperative research report of PWRI (Public Works ResearchInstitute)., The Cooperative research report concerning development ofmethod of observing water level of the levee management technology
upgrade, ” The Water level of the levees observation manual” No.377,
2008.3.
Yukiko Saito, Kazushi Furumoto, Hitoshi Taninaka, Hidetoshi
Kohashi,” Influence of longitudinal seepage flow on stability of river
Levee”, Proceedings of the 62th Annual Conference of JSCE (Japan
Society of Civil Engineers) 3-284, 2007.9.
Yukiko Saito, Hitoshi Taninaka, Hidetoshi Kohashi, Kazushi
Furumoto, ”Influence of longitudinal soil profile on stability of river
Levee” River technological thesis collection vol.14, pp.79-84,
2008.6. ,press release by JSCE
Yukiko Saito, Hirotoshi Mori, Satoshi Arakane, Tetsuya Sasaki, ”The
Influence of three-dimensional seepage flow to the performance ofcountermeasures for river levees”, River technological thesis collection
vol.16, pp.329-334, 2010.6. ,press release by JSCE
Hiroyuki Masuyama, Yukiko Saito, Hirotoshi Mori, Tetsuya
Sasaki, ”On-site monitoring and three-dimensional seepage flow
analysis on the effect of countermeasures for river levees”, River
technological thesis collection vol.17, pp.275-280,2011.7. ,press
release by JSCE
Japan Institute of Construction Engineering (JICE) , ”Technologicalstandard of structure investigation of river leveest” 2002.7

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Influence of Decreasing of Soil Suction on Unconfined Compressive Strength for
Bentonite
T. Nishimura., Department of Civil Engineering, Ashikaga Institute of Technology, JapanABSTRACT: This study focuses on shear strength of compacted
bentonite at different suction values. The soil used in this test
program is sodium bentonite mixed with silica sand. Soil-water
characteristic curve was measured using vapor pressure techniquewhich can provide a relationship between high soil suctions andvolumetric water content. Measured soil-water characteristic curves
were observed few effects of compaction effort. Unconfinedcompression testing on the compacted bentonite with soil suctions
higher than 2.8 MPa were conducted out. Vapor pressure technique
was useful to apply high soil suctions to all compacted bentonite.
Stress and strain curves were measured for various soil suctions, andwere indicated the changing of shear properties due to soil suctions.
Relationship between soil suction and shear strength were evidence
in a linear failure envelope. In addition, influence of rate of axial
strain on unconfined compressive strength was observed in thisstudy.

Keywords: Bentonite, Soil suction, Vapor pressure technique,
Unconfined compressive strength, Relative humidity.

1.5 MPa.
Compacted bentonite with high soil suction
induced the reduction of shear strength due to changing ofrelative humidity without water seepage. Therefore, it isimportant to predict to decrease both stiffness and shear
strength after reduction of soil suction. There are lacks tostudy shear properties of compacted bentonite subjected tochange of relative humidity.
This study focuses on the influence of decreasing of soil
suction on shear strength of compacted bentonite. Before
performing of unconfined compression test, all compacted
bentonite specimens were placed in different relative
humidity conditions using vapor pressure technique. As
results, the different soil suctions were applied to specimens,
and were measured the unconfined compressive strength.
This study presents the relationship between unconfinedcompressive strength and soil suctions.
2. TESTING PROGRAM

1. INTRODUCTION
It is knowledge that bentonite consist of artificial barrier forradioactive waste disposal system. Highly swelling behavior
of compacted bentonite is useful to maintain safety for
geo-environment.
Geotechnicalengineers
andgeo-environmental engineers require interpreting mechanicalproperties of compacted bentonite. Many experimental
papers were published regard to some futures (i.e., thermo
conductivity, chemical properties, hydro-conductivity,
mineralogical composition and swelling behavior). On theother hands, yield surface and constitutive models for
expansive soil as like to compacted bentonite were suggested
in high soil suction ranges with experimental test results.
Osmotic technique or vapor pressure technique was

effectively for applying of high soil suction more than that ofFig. 1. Grain size distribution for silica sand.

2.1 Soil material
Sodium bentonite was used for this test program. Silica sandwas mixed into the bentonite with a ratio of 30 % by dry
weight. The silica sand had a grain size distribution as shownin Fig. 1. The specimen was statically compacted in the steel
mold. The mold had a height of 30 cm and a diameter of 5
cm, respectively. Also, the inside of mold was treated inorder to decreasing the friction. The soil specimen had a dry
density of 1.6 g/cm3 as target value.
2.2 Unconfined compression apparatus
This testing program used a modified triaxial compression
apparatus. The basic construction of a modified triaxial
compression apparatus is similar to the equipment suggested
by Nishimura and Fredlund (2003). The modified triaxial
compression apparatus consisted mainly of a triaxial chamber,a pedestal, a steel mould, a double glass burette, a differential
pressure transducer, a difference displacement sensor, loadcell sensor and relative humidity control circulation system.
The relative humidity control circulation system wasestablished using a conventional pump, along with a smallchamber with salt solution. The air flow maintained aconstant relative humidity surrounding the compacted
bentonite.
2.3 Vapor pressure technique
Soil suction theoretically has a range from 0 kPa to 106 kPa
Table 1. Relationship between relative humidity (RH) andsuction values for different salt solutions at 20 degrees(Delage et al., 1998).

103

Salt solutions(Chemical symbol)
K2S04
KNO3
NH4H2PO4
NaCl
Mg(NO3)2・6H2O
MgCl2・6H2O
LiCl

Relative
humidity (%)
98
95
93.1
75
54
33
11

Suction (kPa)
2,830
6,940
9,800
39,00083,400148,000
296,000

which consists of matric suction and osmotic suction.
Vanapalli et al., (1999) indicated that a suction value overthan 1500 kPa can be considered as high suction. The low
suction (< 1500 kPa) can be controlled in a pressure plateapparatus using a porous plate or a micro porous membrane.
The conventional shear testing apparatus is modified throughthe use of the axis translation technique. The axis translation
technique consists in increasing the ambient air pressure tovalues greater than atmospheric, so as to translate thepore-water pressure into the positive range. In particular axis
translation technique may be not suitable for investigating
unsaturated soils at high degree of saturation because of the
discontinuity of the air phase. The high suction can becontrolled using a vapor pressure technique. High suction
values can be achieved by controlling relative humidity (RH)
using several different salt solutions in a desiccator (Delage etal., 1998). The salt solution method was used in this study toachieve high suction values as summarized in Table 1. The
glass desiccator was then placed in a temperature controlledchamber for at least one month to allow the specimens toachieve an equilibrium condition with respect to thecontrolled suction.

Fig. 2. Applied soil suction .in the glass desiccator.

Fig. 3. Soil-water characteristic curve for bentonite.

2.4 Measurement of soil-water characteristic curve forbentonite
The compacted bentonite specimen with initial water content
from 4.4 % to 5.9 % was placed in glass desiccators withdifferent salt solutions. The seven different salt solutions
summarized in Table 1 were used in the test program. The
gravimetric water content of each soil specimens wasmeasured after achieving the equilibrium condition. As aresult, the relationship between measured gravimetric watercontent and soil suction in high soil suction range can beobtained. The initial samples had a diameter of 2.0 mm and aheight of 1.0 cm, respectively.
2.5 Unconfined compression test
The specimens were statically compacted at initial watercontent from 4.4 % to 5.9 %. The initial soil suction of thespecimen was measured using the soil-water characteristic
curve which gave a suction value of 110 MPa.
The two desired suction values were 148 MPa and 2.8 MPa
corresponding to RH of 33 % and 98 %. Figure. 2 shows the
bentonite specimens were remaining under constant relative

Fig. 4. Soil-water characteristic curve for compacted
bentonite.
humidity in the glass desiccators. At least one month, thebentonite specimens were placed to equilibrium with required
soil suction. The bentonite obviously occurred the swelling
expansion when the soil suction decreased to 2.8 MPa from104

Fig. 5. Volume deformation of compacted bentonite bychanging of soil suction.

Fig. 6. Influence of rate of axial strain on stress and strain
curves.
110 MPa (i.e. initial soil suction). On the contractor, thebentonite specimen was little shrinkage was observed due toincrease of soil suction comparing initial soil suction (i.e. 110
MPa). After ensuring equilibrium conditions, the bentonite
specimen having three differences soil suctions were set up in
the modified triaxial compression apparatus. The specimen
was sheared at a rate of 0.007 % per minute and compression
process was remained to at least 8.0 % in axial strain or until apeak value of deviator stress was observed. For the bentonite
specimen having soil suction of 2.8 MPa three difference
rates were applied in the unconfined compression testing
which were 1 % per minute, 0.013 % per minute and 0.007 %per minute.
3 TEST RESULTS
3.1 Soil-water characteristic curve
Relationship between gravimetric water content orvolumetric water content and soil suction of the bentonite areshown in Figs. 3 and 4. These are difference between two

soil-water characteristic curves in compaction or not.
Changing of gravimetric water content is described against
soil suction because all bentonite samples were no compacted
condition. Other hand, the soil-water characteristic curve in
Fig. 4 was possible to order a volumetric water content thatthe bentonite samples remained a round column in shapethroughout the SWCC testing. Both gravimetric watercontent and volumetric water content decreased withincreasing of soil suctions. The decreasing of moisture inbentonite was large when soil suction regard from 2.8 MPa to
83.4 MPa. Beyond soil suction was 83.4 MPa, the moisturesmoothly decrease with soil suction. Two soil-water
characteristic curves described similar tendency that is few
influence of compaction effort on the soil-water characteristic
curve for bentonite. Figure 5 show relationship between
volume deformation of compacted bentonite and soil suction.
Initial compacted bentonite had a diameter of 2 cm and aheight of 1 cm, respectively. Also, it is predicted that initialsoil suction was 110 MPa. Changing of soil suction induced
the volume change due to vapor hydraulic seepage. The
measured volume strain was over 30 % when soil suction
approached to 2.8 MPa. On the contrary, little shrinkages
were measured when soil suction was over 110 MPa.
Eventually, changing of soil suction gave similar tendency
either volumetric water content or volume strain.
3.2 Stress versus strain curve with various soil suctions
Unconfined compression testing with three difference rate ofshearing was previously conducted using the bentonite
having a soil suction of 2.8 MPa. Applied rates of shearing
were 0.007 %/ per minute, 0.013 % per minute and 1.0 % perminute. Figure 6 shows stress and strain curves. It was notseem to be coincident in three curves. There were large
differences among three test results even maximum deviator
stress. For rate of 0.0 13 % per minute the low stiffness wasobserved comparison to other two stress and strain curves.
Stress and strain curves for compacted bentonite with threedifference soil suctions were determined from unconfinedcompression testing as shown in Figs. 7 to 9. Figure 7 shows
the curve of compacted bentonite having a mixture 30 %when soil suction is 110 MPa. At beginning of shear thedeviator stress increased rapidly, and approached to peckvalue. Beyond the axial strain was 0.77 %, the bentonite
failure that large reduction was described. For soil suction of
148 MPa obtained stress and strain curves were similar to thatof soil suction of 110 MPa. Increasing of deviator stress was
rapidly as shown in Fig. 7. Axial strains at failure had range
from 0.84 % to 1.49 %. Few scatters were observed in three
specimens.
On contractor, the bentonite with soil suction of 2.8 MPa
described the stress and strain curves as shown in Fig. 9. The
deviator stress increased smoothly in shear process. The
bentonite remained the increasing of deviator stress that axial
strain was till 2 %. The deviator stress approached maximum
value when the axial strain was 3.8 % or 4.8 %. After the
bentonite failure, shear resistance described smoothly

105

Table 2. Summary of stiffness with three difference soil
suctions.
Soil suction
(MPa)
2.8
110
148

Stiffness
(MPa)
5.8
100.5
127.9

Fig. 7. Stress and strain curve with soil suction of 110 MPa.

Fig.10. Relationship between unconfined compressive
strength and soil suction.
3.3 Relationship between shear strength and suction
The relationship between unconfined compression strengthand soil suction in the compacted bentonite are shown in Fig.
10. The failure envelope regard to soil suction was linear
relations in range from 2.8 MPa to 148 MPa. The shear
strength of compacted bentonite had effectively influence ofsoil suction, and the angle of friction regard to soil suction
was evaluated as 79.9 degrees.

Fig. 8. Stress and strain curve with soil suction of 148 MPa.

4 CONCLUSIONSThis study focuses on both soil-water characteristic curve andshear strength for compacted bentonite using vapor pressuretechnique. The decreasing of volumetric water content was
obviously large in drying process when soil suction regard
from 2.8 MPa to 83.4 MPa. Beyond soil suction was 83.4
MPa, it smoothly decreased with soil suction. Also,
decreasing of soil suction closely related to reduce the shear
strength for compacted bentonite.
5 REFERENCES[1]

Fig.9. Stress and strain curve with soil suction of 2.8 MPa
reduction. It is clear that unconfined compressive strength forsoil suction of 2.8 MPa was low in comparison with those ofother soil suctions. Also, the stiffness was calculated fromthe stress and strain curves, and summarized in Table 2. It isevidenced that reduction of soil suction closely related todecreasing of the stiffness.

[2]

[3]

106

Nishimura, T. and Fredlund, D.G., “A new triaxial apparatus for hightotal suction using relative humidity control”, 12th Asian Regional
Conference on Soil Mechanics and Geotechnical Engineering, 4-8August Singapore, 2003, pp.65-68
Vanapalli, S.K., Fredlund, D.G. and Pufahl, D.E. ,“The influence ofsoil structure and stress history on the soil-water characteristics of acompacted till.” Geotechnique. Vol.49, No.2, 1999, pp.143-159.
Delage, P., Howat, M.D. and Cui, Y.J., “The relation-ship between
suction and swelling properties in a heavily compacted unsaturated
clay.” Engineering Geology, 50, 1998, pp.31-48.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Relationship between Soil Suction and the Maximum Unsaturated
Undrained Shear Strengths of Compacted Khon Kaen Soil
R. Nuntasarn. Weerawat Wannakul., Department of Civil Engineering, Khon Kaen University



ABSTRACT: The objective of this study was to evaluate themaximum unsaturated undrianed shear strengths of compacted Khon
Kaen soil, which were determined from unconsolidated undrained
triaxial tests. The samples were compacted at 85% of maximum dry
density by standard method. The samples were not measured soil
suction during the test, but rather estimated soil suction from a soil
water characteristic curve (SWCC) based on the volumetric watercontent of samples during the test. The drying soil water
characteristics curve (SWCC) for the compacted Khon Kaen soils
was determined from three methods, which were a hanging column
method, a pressure plate method and an isopiestic humidity method.
The maximum unsaturated undrained shear strength of compacted
Khon Kaen soils was investigated in the transition zone. The
relationship between soil suction and the maximum unsaturated
undrained shear strength of compacted Khon Kaen soil was a good
related in the form of exploration of the value of R2 = 0.98.

Keywords: unconsolidation undrained triaxial test, soil water
characteristic curve, soil suction, maximum unsaturated
undrained shear strength
1. INTRODUCTION
The objective of this study was to evaluate the maximum
unsaturated undrianed shear strengths of compacted Khon
Kaen soil, which were determined from unconsolidated
undrained triaxial tests.
The samples were not measured soil suction during
unconsolidated undrained triaxial tests, but rather estimatedsoil suction from the soil water characteristic curve (SWCC)
based on the volumetric water content of the sample during
the test. The drying soil water characteristics curve (SWCC)
for the compacted Khon Kaen soils was determined fromthree methods, which were hanging column method,pressure plate and isopiestic humidity method.
The volumetric water content during UU test was calculated
from an initial moisture content of sample and volume
change of soil during shear as presented in (1).

θw =Vw w.Ws
V Vi -ΔV(1)

Where w = volumetric water content; Vw = volume ofwater; V = total volume of soil; w = moisture content; Vi= initial volume of specimen; V = volume change of
specimen; and Ws = mass of dry soil.
Due to both of pore water and pore air pressures were notmeasured during unconsolidated undrained triaxial tests.
Therefore the maximum unsaturated undrained shear

strength of compacted Khon Kaen soils was related to thetotal stress.
2. BASIC PROPERTIESThe basic and engineering properties of Khon Kaen soils arepresented in Table 1.
The USCS classifications of Khon Kaen soil based on the
results of soil particle distribution curves and Atterberg
limits.
The maximum dry density and optimum moisture content
Khon Kaen soil was determined by standard compactive
effort (ASTM D698 – 00a).
Table 1. The basic and engineering properties of Khon Kaen
soils
Particle Size Distribution
Clay (%)
13
Silt (%)
31
Sand (%)
56
Atterberg Limit
LL (%)
20.3
PL (%)
14.5
PI (%)
5.8
USCS Classification
SM -SC
Gs
2.65
Standard Compaction
OMC (%)
8.25
Maximum dry density (t/m3)
2.00
3. SOIL WATER CHARACTERISTIC CURVEThe initial dry density, moisture content and void ratio ofcompacted Khon Kaen samples were 1.7 t/m3 (or 85% of
maximum dry density), 14 % and 0.47, respectively.
The drying soil water characteristics curve (SWCC) for thecompacted Khon Kaen soils was determined from threemethods, which were a hanging column method, a pressure
plate and an isopiestic humidity method.
According to [1], the hanging column method was used todetermine suctions in the range of 0 to 9.81 kPa.
The pressure plate method (refer to [2]) was used todetermine SWCC for a suction values between 10 to 1,500
kPa.
Finally, SWCC at suctions above 1500 kPa were determinedfrom the isopiestic humidity method. In this method threesolutions, which were Copper Sulphate (CuSO4), Ammonium
Chloride (NH4Cl) and Sodium Hydroxide (NaOH.H2O) wereused to determine SWCC at 3,900, 30,900 and 365,183 kPa
for a suction value, respectively. Data points above 1500 kPa
were total suction values [3].

107The relationships between soil suction and volumetricmoisture content are illustrated in Fig. 1.
Fig. 1 showed that the projected air entry value for compacted
Khon Kaen soils was equal to 12 kPa and a very low residual
volumetric moisture content of 1.0%. Therefore the suction
range of transition regime was between 12 to 45,000 kPa.

Fig. 1. The relationship between soil suction and volumetricmoisture content of compacted Khon Kaen soil
4. UNSATURATED UNDRAINED SHEAR
STRENGTH
Ten samples of Khon Kaen soils were compacted into amold with a diameter of 50 mm and 100 mm high to obtain
the initial dry density, moisture content and ratio of 1.7 t/m3,
14%, and 0.47, respectively. All samples were determined
an undrained shear strength by the unconsolidated undrained
test (UU test) at a confining pressure of 100 or 200 kPa. Two
samples were tested immediately after compaction. Anothereight samples were air dried for 2, 12, 24 or 72 hours prior toundrained shear. The initial moisture content was measured
after shear to determine the initial soil suction by assumed
the moisture content during testing was not changed.
The volume changes of soil samples were monitored as the
sample was loaded at an axial strain rate of 0.1 % via thevolume of water flowing out of, or into, the triaxial chamber.
The volumetric water content during UU test was calculated
from an initial moisture content of sample and a volume
change of soil during shear as presented in (1).
The summaries of initial test condition and result of UU testfor unsaturated compacted Khon Kaen soils is provided inTable 2 and 3, respectively.
The relationship between stress and strain of unsaturated
compacted Khon Kaen soils for UU test in Fig. 2 showed thestrain hardening behavior, except that soil samples were air
dried for 72 hours prior to shear. Therefore the failure wasdefined as 15% axial strain except that soil samples with 72
hours air dried. The failure of soil samples with 72 hours air
dried was defined as the maximum deviator stress.
The relationship between axial and volumetric strain, as
presented in Fig. 3, indicated that the sample, which was air
dried for 72 hours and subjected to confining pressure of 200
kPa, exhibited dilation. The initial moisture content andvolumetric moisture content of this sample was 3 % and
4 %, respectively, which was corresponded to soil suction of
15,230 kPa.

Table 2. The initial condition of unconsolidated undrained
test for compacted Khon Kaen soils
Initial
Initial soil
SampleInitialmoisture
volumetricsuctioncontent
moisture(kPa)
content (%)(%)
CP =100 kPa
14.9
25
53
(0 hrs-air dry)
CP =200 kPa
14.4
21
188.5
(0 hrs-air dry)
CP =100 kPa
14.3
24
87
(2 hrs-air dry)
CP =200 kPa
13.6
20
245
(2 hrs-air dry)
CP =100 kPa
10.8
18
419
(12 hrs-air dry)
CP =200 kPa
9.6
16
705
(12 hrs-air dry)
CP =100 kPa
8.8
15
1000
(24 hrs-air dry)
CP =200 kPa
7.2
12
2026
(24 hrs-air dry)
CP =100 kPa
4.7
6
9263
(72 hrs-air dry)
CP =200 kPa
3.1
4
15230
(72 hrs-air dry)
Table 3. The summaries of unconsolidated undrained test forcompacted Khon Kaen soils
Samplesoil suction
3f
max
1f
at failure(kPa) (kPa)
(kPa)
(kPa)
CP = 100 kPa
37
100
170
35
(0 hrs-air dry)
CP = 100 kPa
52
101
183
41
(2 hrs-air dry)
CP = 100 kPa
270
98
259
81
(12 hrs-air dry)
CP = 100 kPa
716
98
344
123
(24 hrs-air dry)
CP = 100 kPa
8779
101
1281
590
(72 hrs-air dry)
CP = 200 kPa
118
201
304
52
(0 hrs-air dry)
CP = 200 kPa
149
199
348
75
(2 hrs-air dry)
CP = 200 kPa
460
199
487
145
(12 hrs-air dry)
CP = 200 kPa
1576
198
647
225
(24 hrs-air dry)
CP = 200 kPa
17467
198
1560
681
(72 hrs-air dry)The plot between deviator stress and soil suction for samples
with high soil suction as provided in Fig. 4 indicated that thesoil suction dropped significantly after the peak deviator
stress had been reached as presented in Fig. 2 due to the
behavior of soils as a strain-softening. On the other hand, thestress-strain behavior of samples with low soil suctions108

indicated the strain hardening (see Fig. 2). Therefore the soil
suction remained constant thereafter the residual soil suction
had been reached as presented at Fig. 5.
Fig. 6 and 7 presented that the soil suction of unsaturated
compacted Khon Kaen soils with a high initial soil suction
value was deceased continuously with axial strained.
Meanwhile the soil suction of the other soil samples wasquite constant with axial strain.
5 DISCUSSION
The soil suction range, which used in this experiment, wasbetween 35 to 17,500 kPa. This range of soil suction was on
the transition zone. This indicated that the samples were
unsaturated. According to [4] the soil suction range between
100 to 10,000 kPa is on the adsorbed film regime. The
bonding between water and the soil particle on this regime isshort-range solid-liquid interaction mechanism. A soil
suction value of the samples, almost all is in the adsorbed film
regime. This explained that the bonding between water andsoil particle of the compacted Khon Kaen soils was
short-range solid-liquid interaction mechanism.
Fig. 8 showed the extended Mohr circles at failure as total
confining pressure of 100 and 200 kPa. The typical Mohr
circles at failure of compacted Khon Kaen soils as total
confining pressure of 100 and 200 kPa was presented in Fig. 9
and 10, respectively. Fig. 9 and 10 was illustrated the total
stress path of various soil suctions. Then the plot of maximum
shear versus soil suction on a log scale as 100 and 200 kPa
total confining pressure was illustrated in Fig. 11 and 12,
respectively. Fig. 11 and 12 was provided the path of soil
suction in term of total stress. These paths showed a good
non-linearly correlation between the maximum undrained
shear strength and soil suction. However, a soil suction valueof this study was not measured but it was estimated from
SWCC.

Fig. 3. The relationship between vertical and volumetricstrain of unsaturated compacted Khon Kaen soils from UU
testing

Fig. 4. The relationship between a high suction and deviator
stress of unsaturated compacted Khon Kaen soils from UU
testing

Fig. 2. The stress-strain curve of unsaturated compacted
Khon Kaen soils from UU testingFig. 5. The relationship between a low suction and deviator
stress of unsaturated compacted Khon Kaen soil from UU
testing109

Fig. 6. The relationship between a high suction and axial
strain of unsaturated compacted Khon Kaen soils from UU
testing

Fig. 10. Typical Mohr circles at failure with various soil
suctions for compacted Khon Kaen soils as 200 total
confining pressures

Fig. 7. The relationship between a low suction and axial strainof unsaturated compacted Khon Kaen soils from UU testingFig. 11. The soil suction path of compacted Khon Kaen loessas 100 kPa total confining pressure

Fig. 12. The soil suction path of compacted Khon Kaen loessas 200 kPa total confining pressure

Fig. 8. The extended Mohr circles of compacted Khon Kaenloess

6 CONCLUSION

Fig. 9. Typical Mohr circles at failure with various soil
suctions for compacted Khon Kaen soils as 100 total
confining pressures

Khon Kaen loess was compacted at 85% of maximum dry
density by standard compactive effort to determine theunsaturated undrained shear strength parameter from theunconsolidated undrained test. Due to the pore water andpore air pressure did not measure during test, since the soil
suction was estimated from SWCC.
The SWCC of compacted Khon Kaen loess showed theresidual volumetric moisture content of 1.0 % and the air
entry value of 12 kPa.
Finally, the path of soil suction during undrained
unconsolidation test in term of total stress showed a good110

non-linearly correlation between the maximum undrained
shear strength and soil suction.
7 ACKNOWLEDGMENT
Acknowledgement is given to Faculty of Engineering, Khon
Kaen University for the support of this re-search. Sincere
thanks are extended to Dr. Ketsuda Dejbhimon and Mrs
Pornpis Chuson, for the favor of using instruments in thisstudy. Special thanks to Ms. Nichapat Banreangthong,
Mr.Tanakorn Dua-ngkantai, Mr. Nanthiwa Soontararuk, Mr.
Piyanat Jantosut, Mr. Veeraphon Tansa, Ms. Pemiga
Thong-tan, and everyone who have been involved directly
and indirectly in the preparation of this research.
8

REFERENCES[1]

ASTM - American Society for Testing Material, ‘Standard Test
Methods for Determination of the Soil Water Characteristic Curve for
Desorption Using a Hanging Column, Pressure Extractor, Chilled
Mirror Hygrometer, and/or Centrifuge1’, (ASTM D6836 – 02), WestConshohocken, PA., 2002
ASTM - American Society for Testing Material, ‘Standard Test
Method for Capillary-Moisture Relationships for Coarse- And
Medium-Textured Soils by Porous-Plate Apparatus’, (ASTM
D2325-98), West Conshohocken, PA., 1998
Lu, N. and Likos, W.J., Unsaturated Soil Mechanics, John Wiley &Sons, Inc., Hoboken, New Jersey., 2004, ch. 10
McQueen, I.S. and Miller, R.F., ‘Calibration and Evaluation ofWide-Range Gravimetric Method for Measuring Soil MoistureStress’, Soil Science, 10(3), 1968, pp 521-527.

[2]

[3]
[4]

111

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Interpretation of Coaxial and Non-Coaxial Strain on Directional Variation of
Undrained Strength of clay
J.Rajaraman1 and K. Thiruvenkatasamy2



ABSTRACT: The term stresses on soils, soil samples are muchdiscussed by soil researchers , but the strains are not. Stresses areinvisible ,but the strains [deformations] are visible. Strain describes
the distortion of a body in response to an applied force. Strain ishomogeneous when any two portions of the body that were similar
in form and orientation before are similar in form and orientation
after strain. Homogeneous by definition means the strain ishomogeneous. Conversely, heterogeneous strain implies
heterogeneous deformation. Because heterogeneous strain is more
complex to describe than homogeneous strain, evenheterogeneously strained bodies or regions are analyzed byseparating
them
into homogeneous portions. In other words,
homogeneity of deformation is a matter of scale.
In this paper the directional variation of undrained strength of clay
as identified by Casagrande and Corrillo [1]. [3]. is taken as
illustration and interpreted in terms of Coaxial and Non coaxial
strain as analytical tool.

1. INTRODUCTION
Isotropic stress acts equally in all directions, it results in a
volume change of the body.Deviatoric stress, on the other
hand, changes the shape of a body [2]material lines form the axes of an ellipse that is called strain
ellipse.The principal incremental strain axes rotate to thefinite strain axes, a scenario that is called non- coaxial strain
accumulation.The case in which the same material lines
remain the principal strain axes at each increment is called
coaxial strain accumulation.
2.0 Undrained Shear Strength of Clay
Let the undrained shear strength of a soil specimen with its
axis vertical be referred to Su[V]. Similarly let the undrained
shear strength with its axis horizontal be referred to as Su[H],If Su[V]. = Su[i] = Su[H] the soil is isotropic. Here Su{i} is theundrained shear strength when he major principal stressmakes an angle i with the horizontal. If the soil is anisotropic,
Su[i] will change with direction. Cassagrande and Corrillo
proposed the following equation for the directional variation
the undrained shear strength; Su[i]= Su[H] + [Su[V]-Su[H] ] sin2 i. For this illustration the figures and calculations
are transformed into Coaxial and Non-coaxial diagrams withminimum changes. The Circle and the strained ellipsoids
help to identify the anisotropic behavior of clay. Since thesample is pure clay only pure shear is involved and and thesimple shear is applicable only for sand .The following
figure 2 explains simple shear and pure shear . [2]Figure 1
Figure 1

Figure 1 The mean (hydrostatic ) and deviatoriccomponents of the stress. (a)mean stresscauses volume
change and (b) Deviatoric stress causes shape change
In a homogeneously strained, two-dimensional body there
will be at least two material lines that do not rotate relative toeach other, meaning that their angle remains the same before
and after strain. A material line connects features , such as anarray of grains, that are recognizable throughout a body’s
strain history. The circle deforms and changes into an ellipse.
In homogeneous strain, two orientations of material lines
remain perpendicular before and after strain. These two

Figure 2 The rigid spheres slide past one another to
Accommodate the shape change without distortion of
the individual marbles . In figure 2b the shape changeis achieved by changes in the shape of individual clay
balls to ellipsoids, are quite different.

Figure 222

2. VARIATION OF UNDRAINEDSHEAR STRENGTH
In this paper the directional variation of undrained strengthofclay as identified by as Casagrande and Corrillo [1] .[3]. is
1,2 Professor, Department of Harbour Engineering and Off Shoretaken as illustration and interpreted in terms of Coaxial andTechnology , AMET University, 135 East Coast road, Kanathur -603 112 ,Non coaxial strain as analytical tool.
CHENNAI, INDIAO
Anisotropy in undrained shear strength:
E-Mail rajaraman.usha44@gmail.com1
Owing to the nature of the deposition of cohesive soils andswamy2667@yahoo.co.in2
113subsequent consolidation, clay particles tend to becomeoriented perpendicular to the direction of the major
principal stress.

Figure 4 Directional Variation of undrained strength of clay.
If Su(V) =Su(i) =Su(H), the soil is isotropic with respect to strength,
and the variation of undrained shear strength can be represented
by a circle in a polar diagrams shown by curve a in figure 4.

If

If the soil is anisotropic, Su(i) will change with direction.
Casagrande and Carrillo (1944) proposed the following
equation for the directional variation of the undrained shear
strength ;(2)Su(i) = Su(H) +[Su(V) – Su(H) ] sin2 i
4. COAXIAL AND NON-COAXIAL STRAIN ANALYSIS

Figure 3 Strength anisotropy in clay.
Parallel orientation of clay particles could cause the strength ofthe clay to vary with direction, or in other words the clay could
be anisotropy with respect to strength. This can bedemonstrated with the aid of the above figure, in which V, H
are vertical and horizontal directions that coincide with lines
perpendicular and parallel to the bedding planes of a soil
deposit.
If a soil specimen with its axis inclined at an angle i with thehorizontal is collected and subjected to an undrained test,
the undrained shear strength can be given by(1)
Su(i) = σ1-σ3

Figure 9 Homogeneous strain describes thetransformation of a square to a rectangle ora circle to an ellipse. Two material lines thatremain perpendicular before and after strain
are the principal axes of the strain ellipse [solid lines].
The dashed lines are material lines that do not
remain perpendicular after strain; they rotate
toward the long axis of the strain ellipse.

2

Where Su(i) is the undrained shear strength when the major

principal stress makes an angle i with the horizontal.
Let the undrained shear strength of a soil specimen with its
Axis vertical [i.e; Su(i=90o)] be referred to as Su(V) Figure 3(a).
Similarly the undrained shear strength with its axis
horizontal [i.e; Su(i=0o)] be referred to as Su(H) figure 3(c).

114

Figure10 A combination of simple shear [a special case ofnon-coaxial strain] and pure shear [coaxial strain]is called
general shear or general non-coaxial strain. Two types ofgeneralshear
aretranstension[a]andtranspression[b],reflecting extension and shortening
components.

When σ1 acts parallel to the bedding plane as compressive
axial stress and i =90o the major principal stress σ1 makes anangle 90o with the vertical. In this case the circle becomes
ellipse elongated in y direction and compressed in x direction.
This is similar to curve b in figure 4.
This circle representing the curve a in figure 4 and it takes apath in between the deformed ellipse representing
the directional variation of the shear strength. The combined
diagram of figures 5, 6 &7 is given below.

5. COAXIAL STRAIN INTERPRETATION
Now the discussion can be extended to co-axial strain
analysis .The following figures will be useful forinterpretations.

Figure 7

Figure 5.
When i =0 the major principal stress makes an angle 900 withthe horizontal plane(bedding plane ) as identified in figure
3b.Since σ1 is compressive stress acting vertically down , thecircle becomes ellipse elongated along x direction andcompressed in y direction. This is similar to the curve cshown in figure 4

Figure 8 Co-Axial strain on clay ( pure Shear )5.1 Coaxial Strain and Anisotropy
When Su(V) > Su(H) , the nature of variation of Su(i) can be
represented by curve b in figure 2. If Su(V)< Su(H) , thevariation of Su(i) Is given by curve c .The coefficient of
anisotropy can be defined as :K= Su(V)
Su(H)

Figure 6115

2. Figure 4 is confined to quadrant 1only.Figure 8
covers all 4 quadrants. And includes figure 4 (curveb,a and c).
3. Strain ratios measured agree with the the familiar
coefficient of anisotropy ( K) which ranges from0.75 to 2.0 (as shown in Table I)
4. X axis and Y axis form material lines perpendicular
to each other. As a result the deformation X/Y or
Y/X as strain ratio are calculated
5. i and strain ratios are related as shown in Table II

In the case of natural soil deposits , the value of K can varyfrom 0.75 to 2.0. K is generally less than 1 inoverconsolidated clays.

TABLE I Coaxial Strain And Anisotropy (K)TABLE II Shear Strength and Coaxial Strain Ellipses

i = angle with the horizontal plane (bedding plane )MPS – major principal stress
σ1, σ3 are major and minor principal stresses.
6．CONCLUSION :1. Pure clay is associated with pure shear or coaxial
strain .No simple shear.
７．REFERENCES
116

1.Casagrande A. and N. Carrillo, Shear failure ofanisotropic materials, in contribution to soil
mechanics 1941-1953 , Boston society of Civil
engineers , Boston , Mass ., 1944
2. Van der pluijm, ben a, Earth Structure-2003
W.W.Norton & company inc, New York 10110.
3.Das Braja M,2008Advanced Soil Mechanic, Taylor& Francis pp.433-435.

117

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

FEM Analysis of Innovative Shallow Foundation System for Reactive Soils
Mostafa A. Ismail, Advanced Geomechanics, Perth WA, Australia
Mohamed A. Shahin, Curtin University, Department of Civil Engineering, Perth WA, Australia


ABSTRACT: Granular pile anchor foundations (GPAF) are apromising foundation system that can be used to mitigate the serious
consequences of volumetric changes of reactive soils, both during
expansion and shrinkage. This paper presents results from 3D finite
element analyses, using PLAXIS software, undertaken on a typical
double-story building constructed over a system of GPAF in areactive soil. The study investigates the ability of the GPAF system
to resist the forces induced by soil movement due to moisturevariation, and the impact of this resistance on the straining actions
affecting the superstructure. The results confirm the efficiency of the
GPAF system in arresting the movement of the reactive soil, which
in turn improves the structural responses of the building in terms of
induced deformations, angular distortions and internal forces.

1. INTRODUCTION
Reactive soils are clays that swell and shrink with changing
moisture content, and are found in many arid and semiarid
regions around the world. For example, it is estimated that
20% of the Australian surface soil can be classified as reactive[1]. There are several factors that influence the swelling andshrinking potential of clay soils, including the amount andtype of clay minerals, cation exchange capacities, availabilityof moisture and the initial water content. The swell/shrink
ground movements associated with reactive soils posesignificant challenge to the geotechnical community as theycause distress to foundations of lightweight structures andcracking in retaining walls, pavements, canal beds and linings[2, 3].
Despite the numerous foundation systems that have beendeveloped over the past decades to control the movements
induced by reactive soils, substantial financial losses are still
incurred every year in many places around the world. For
example, the American Society of Civil Engineers estimatedthat about one quarter of all homes in the US haveexperienced some damage from reactive soils; the financial
losses incurred by property owners exceed those caused bynatural disasters such as earthquakes, floods, hurricanes andtornadoes combined [4]. In Australia, and despite thestringent regulatory requirements, most of the lightweightbuildings constructed on reactive soils experience some
distortional damage during their early lives [5].

promising, special foundation solution that has been recently
proposed is the granular pile anchor foundation (GPAF). The
GPFA is an innovative technique that has shown great
potential as a solution to founding structures safely into
reactive soils. The GPAF system was first proposed by
Phanikumar and Ramachandra Rao [9] for reactive soils
under heave conditions and was later pursued by otherinvestigators [e.g. 10, 11] via laboratory and field trials. In an
attempt to determine the controlling parameters of the GPAF
technique to resist the heave and shrinkage induced byreactive soils, the efficiency of the technique was further
investigated numerically by Ismail and Shahin [12] using thefinite element method (FEM). Despite the success of the
GPAF reported by the above investigators, it is yet to be
applied in practice, primarily due to the limited field trials.
In this paper, the performance of the GPAF system underheave and shrinkage is investigated by 3D numerical analysis
for a typical double-story, four-bay structure founded inreactive soil. The investigation explored the efficiency of thesystem in arresting the ground movements induced by soil
heave/shrinkage, and the implications of this on the internal
forces experienced by the superstructure.
2. CONCEPT OF GPAF SYSETM
Fig. 1 shows the concept of the GPAF system, which is ahybrid solution in which a shallow foundation is supported ona granular pile that derives its resistance from the interface
between the granular pile and surrounding reactive soil.

The literature includes numerous solutions for construction
on reactive soils, including replacement of the entire reactive
material, pile foundations [6], soil stabilization using
additives [7, 8], and implementation of special types offoundations such as drilled and friction piers [3]. One

117

Pile Anchor

Footing

Heave Uplift Pressure
Pile Uplift
Resistance

Granular Pile

Expansive Soil
Anchor Plate

Fig. 1: Typical GPAF systemdisplacement nodes and 6 Gaussian stress points each.
Pad footing
0.5 m
0.1 m

2m

As can be seen in Fig. 1, a typical GPAF system consists of apile of granular material installed into the reactive soil. A
concrete footing is then constructed above the granular pileand connected to it via a steel anchor that is monolithicallycasted with the concrete footing to transfer the load between
the footing and the pile. The uplift resistance is mobilized asskin friction along the periphery of the granular pile at its
interface with the surrounding soil. The force in the pileanchor is transmitted to this interface via a base plate that is
rigidly connected to the pile anchor. According to thisarrangement of the anchorage system between the footing andgranular pile, the latter cannot only reinforce the ground (as in
the case of soft clay and loose sand) but can also effectively
resist the uplift forces from an expansive soil. As can beinferred from Fig. 1, the uplift resistance of the GPAF system
is a function of the self-weight of the pile-footing assembly,interface shear strength, surface area of the granular pile andnormal stress developed during expansion of the soil
surrounding the pile. In the experiment performed by
Phanikumar et al. [13], the undrained shear strength of the
expansive clay surrounding the granular pile was shown tohave increased by about 20%, compared with the clay in thefree field zone due to the expansion-induced normal stresses.

Granular pile
Pile anchor
2mFig. 2: Schematic diagram of the pad footing with three
granular pile anchors

Foundation Level(a)

3. NUMERICAL ANALYSES OF TWO-STORYBUILDING RESTING ON GPAF SYSTEM
In order to investigate the efficiency of the GPAF system inpractice, a two-story four-bay frame building resting on asystem of GPAF is considered in the current study. The
problem is analyzed by numerical modeling using thecommercially available finite element software PLAXIS 3DFoundation [14].

Ground beam
Heave/Shrinkage area(b)

3.1 Problem Identification
The two-story frame building considered in the current study
is 6 m high (each story is 3 m in height), and is 20 m × 20 m inplan with each bay having dimensions of 5 m × 5 m. A ceilingslab of 160 mm thick is assumed for each story. The slabs aresupported by beams, 300 mm wide and 400 mm deep, which
in turn rest on square columns of dimensions 300 mm × 300
mm. The dead load of each structural component of the framebuilding is considered according to the material unit weight ofthat component, and an additional distributed live load of 5
kPa is also assumed to act on top of the slabs. All concretematerials (including footings) are made of concrete of anelastic modulus of 35 GPa, Poisson’s ratio of 0.2 and unit
weight of 24 kN/m3.
The GPAF system consists of square pad footings ofdimensions 2 m × 2 m, each supported on a group of three
granular piles of 0.5 m in diameter and 3.0 m in length. A
schematic diagram of the arrangement of the three granularpiles within the pad footing is shown in Fig. 2. A group ofpiles is used rather than a single pile to enhance the rotational
stiffness and stability of the system. The problem is analyzed
using the 3D model presented in Fig. 3, which has adiscretized mesh that consists of 17,880 wedge elements of 15

Footing(c)Fig. 3: FEM 3D model: (a) double-story building including
GPAF system; (b) plan view of building foundation; (c) meshdiscretization

118

The idealized ground profile consists of a gently sloping
reactive clay layer of an average thickness of 3.5 m (acrossthe footing foot print) overlying dense sand of an average
thickness of 11.5 m. The model is strategically refined aroundthe footing and granular piles to improve the accuracy of theanalysis and the boundaries are located farther from the areaof interest to minimize the boundary effect. The concrete
footings are located 1.0 m below the ground surface and aremodeled using a Mindlin’s plate element of thickness of 0.6
m. The pile anchor is modeled as an elastic embedded pile[14] of 75 mm in diameter with an elastic modulus of 200 GP.
The steel base plate is modeled as an infinitely rigid plate. All
pad footings are connected by concrete tie beams (see Fig. 3b),
300 mm wide and 600 mm deep.
3.2 Soil Models and Parameters
The reactive clay is modeled using a Mohr-Coulomb (MC)model and is assumed to behave in a drained manner during
expansion and shrinking. The underlying dense sand andgranular pile material are best modeled using the hardening
soil constitutive model (HS). The HS model [15] is a
non-linear elastic plastic formulation which adopts multiple
yield loci as a function of plastic shear strain and a cap toallow volumetric hardening. The non-linear stress strain
relationship is represented by a hyperbolic formula, withprimary loading governed by a secant deformation modulus(E50) at 50% of the material strength. Loading and unloading
within the current yield surface are assumed to be elastic(defined by a separate modulus, Eur) with failure governed by
the Mohr-Coulomb failure criterion. Both E50 and Eur evolvewith the minor effective stress,  ' 3 , according to thefollowing formula: c cos   3 sin  E 50  E50ref  c cos  p sin  
ref



m

(1)

shrinkage events are applied independent of each other,starting from the stage after application of the dead and live
loads.
Table 1. Soil properties used in the finite element analyses
Soil Type
Dense
Sand
20

GranularPile
20

E 50ref (MPa)

2

75

200

ref
oed

–

75

200

ref
ur

̶
2
24
0.35
–
–
0.6

200
0.1
36
0.2
100
0.5
0.4

600
0.1
40
0.2
100
0.5
0.4

E (MPa)E (MPa)c (kPa) (°)

ur( 2)pref (kPa)m
K0(3)(1) Soil unit weight; (2) Unload-reload Poisson’s ratio; (3) Coefficient ofearth pressure at rest.

4. RESULTS AND DISCUSSION
In order to investigate the efficiency of the GPAF system inenhancing the behavior of the two-story building constructed
over a reactive soil, an additional independent analysis iscarried out for the building resting on pad footings without
the GPAF and the results are compared with those obtained
from the analysis of the building resting on the GPAF system.
The comparison is made for the top beams of the central
frame (i.e. beams B1 to B4 in Fig. 4). The comparison resultsin terms of induced deformations due to heave and shrinkage
are shown in Fig. 5, and the angular distortions and bending
moments due to heave are given in Table 2.
B1

B2

B3

B4

5m

5m

5m

5m

3m

3m

where: c is the soil effective cohesion, ϕ is the effective peakfriction angle, m is the exponent that controls the dependencyof the stiffness on stress and pref is the reference stresscorresponding to E 50ref . A summary of the material parameters
used for all soils are presented in Table 1. It should be notedthat the properties of the reactive clay are those evolving after
the wetting/drying event and during expansion/shrinking. In
reality, the strength of a reactive soil degrades during
expansion and increases during shrinking (due to suction), but
this is not modeled in this study.
The rate of volume change at which reactive clays would
normally encounter depends on the location from the sourceof moisture and magnitude of overburden pressure. In thecurrent study, a leaking event of an underground water
facility located underneath the central column is assumed tocause arbitrary values of heave and shrinkage of 20% and10%, respectively, over the layer thickness of the affectedarea underneath the central footing, as shown in Fig. 3b. Bothheave and shrinkage were modeled by applying equivalent
volumetric strains to the affected area. The heave and (kN/m3)(1)

Reactive
Clay
15

Parameter

3m

1m

GS

Footing
Reactive clay

Dense sandFig. 4: Schematic diagram of the central frame
It can be seen from Fig. 5 that the ability of the pad footings toresist the vertical movements induced by soil heave issignificantly improved when the GPAF system is used. The
maximum vertical displacement induced by soil heave for119beams B1–B4 without the GPAF system is found to be equal
to 6.7 mm, whereas negligible vertical movement isdeveloped when the GPFA system is used. More importantly,
it can be seen from Table 2 that all beams undergo much lessangular distortions when the GPFA system is used. For
example, the angular distortion of beam B2 without the GPAF
system is some 300 times greater than that experienced when
the GPAF system is used.
8
Heave 20% - GPAF

As can be seen in Table 2, the use of the GPAF systemsignificantly reduced the maximum negative bending
moments of all beams, but slightly reduced the maximum
positive bending moments. For example, the maximum
negative and positive moments of beam B1 are 81.2 and 29.3
kN.m, respectively, for the case without the GPAF system,
whereas these values are reduced to 44.6 and 23.0 kN.m,
respectively, for the case with the GPAF system.

Heave 20% - No GPAF

6

Shrinkage 10% - GPAF

Beam deformation (mm)

Shrinkage 10% - No GPAF

42

Beam Initial Position

0-2

The practical implication of the above results is that the use of
the GPAF system for light-weight structures can significantly
reduce the superstructure damage induced by reactive soils,
leading to immense savings on the cost of structural repairs
and ongoing maintenance.

-4
-6
-10

-5

0

5

capacity of the sand layer that embrace the base plate (there
are evidences that down drag forces are transferred to theshrinking soil). Given that the granular pile has no tension or
bending capacity, it is therefore expected that the maximum
capacity to resist shrinkage will be reached when theshrinking soil detaches itself completely from the granularpile. In such case the shrinkage resistance can be significantly
improved by encasing the granular pile into a stiff, geogrid
case to stop the pile from bulging.

10

Ditance from beam centerline (m)5. SUMMARY AND CONCUSIONS

Fig. 5: Deformation of central beams B1 to B4 due toheave and shrinkage

This paper presented results from 3D FEM analyses of the
granular pile-anchor foundation (GPAF) system as aplausible foundation solution for light-weight structures built
on reactive soils. The paper investigated the superstructure
response of two-story, four-bay frame building to the groundheave/shrinkage induced by reactive soils.

Table 2. Angular distortions and internal forces of central
beams B1 to B4 due to heave
Beam
Number

B1
B2
B3
B4

GPAF
No GPAFGPAF
No GPAFGPAF
No GPAFGPAF
No GPAF

Angulardistortion(× 10–5)
0.6
50
0.2
59
2.8
64
1.8
51

Maximum
negative
moment(kN.m)
44.6
81.2
59.8
130.2
58.0
132.4
44.0
79.7

Maximum
positive
moment(kN.m)
23.0
29.3
26.0
25.3
28.0
28.6
25.0
29.6

Note: 20% Heave: Free field heave = 75 mm

Previous investigations found in the literature did not address
the suitability of the GPAF system to resist shrinkage events
when reactive soil loses moisture. However, it can be readily
shown from Fig. 5 that under such event the GPAF system iscapable of arresting the shrinkage and significantly reducing
its induced settlement. It is found that the maximum beam
settlement induced by soil shrinkage for the building withGPAF system is reduced by about 75% compared with thecase of no GPAF system. It should be noted that the capacityof the GPAF system to resist shrinkage is a result of its ability
to bear directly on the piles (while still in contact with theshrinking soil), which in turn could engage the bearing

The results indicate that the effect of the GPAF system in
reducing the vertical displacement and angular distortion
induced by soil heave are quite significant. It was shown thatthe maximum vertical displacement of 6.7 mm that isdeveloped at the top beams of the central frame in the case of
no GPAF has been totally arrested by the GPAF system, andno heave induced vertical displacement is produced. It wasalso shown that the GPAF system reduced the maximum
angular distortion of the top beams dramatically. In
accordance with this, the GPAF system is found to reduce themaximum negative and positive bending moments of the top
beams by 56% and 21%, respectively, compared with the
maximum negative and positive bending moments producedin the case of no GPAF system. It was also observed that theresistance to shrinkage is improved immensely when the
GPAF system is used. The maximum settlement induced byshrinkage for the top beams of the central frame is reduced by
75% compared with the case of no GPAF system.
The above results conclude that the GPAF system is apromising foundation technique that can be potentially used
to reduce the detrimental impacts of reactive soils under bothheave and shrinkage conditions.

120

6. REFERENCES
[1]
[2]
[3]
[4]
[5]

[6]
[7]
[8]

Richards BG, Peter P, and Emerson WW, "The effects
of vegetation on the swelling and shrinking of soils inAustralia," Geotechnique, 33(2), 1983, pp. 127-139.
Al-Rawas A, and Goosen MFA, Reactive soils, recent
advances in characterization and treatment. London:
Taylor & Francis, 2006.
Chen FH, Foundations on reactive soils. New York:
Elsevier, 1988.
Wray WK, So your home is built on expansive soils: a
discussion of how expansive soils affect buildings. New
York: American Society of Civil Engineers, 1995.
Barthur R, Jaksa MB, and Mitchell PW, "Design ofresidential footings built on expansive soil using
probabilistic methods," in Proceedings of the 7thAustralia New Zealand Conference on Geomechanics,
1996, Adelaide, pp. 369-374.
Nusier OK, Alawneh AS, and Abdullatit BM,"Small-scale micropiles to control heave on reactive
clays," Ground Improvement, 162(1), 2009, pp. 27-35.
Ene E, and Okagbue C, "Some basic geotechnicalproperties of reactive soil modified using pyroclastic
dust," Engineering Geology, 107(1-2), 2009, pp. 61-65.
Khattab SAA, Al-Mukhtar M, and Fleureau JM,"Long-term stability characteristics of a lime-treatedplastic soil," Journal of Materials in Civil Engineering,19(4), 2007, pp. 358-366.

[9]

Phanikumar BR, and Ramachandra Rao N, "Increasing
pull-out capacity of granular pile anchors in reactive
soils using base geosynthetics," Canadian Geotechnical
Journal, 37(4), 2000, pp. 870-881.
[10] Phanikumar BR, Srirama Rao A, and Suresh K, "Field
behaviour of granular pile-anchors in reactive soils,"
Ground Improvement, 161(G14), 2008, pp. 199-206.
[11] Sharma RS, Phanikumar BR, and Nagendra G,"Compressive load response of granular piles reinforced
with geogrids," Canadian Geotechnical Journal, 41(1),2004, pp. 187-192.
[12] Ismail AM, and Shahin MA, "Finite element analyses of
granular pile anchors as a foundation option for reactive
soils," in International Conference on Advances inGeotechnical Engineering, 2011, Perth, Western
Australia, in press.
[13] Phanikumar BR, et al., "Granular pile anchor foundation(GPAF) system for improving the engineering behaviorof expansive clay beds " Geotechnical Testing Journal,
27(3), 2004, pp. 1-9.
[14] Brinkgreve RB, and Broere W, PLAXIS 3D foundation
manual, Version 2.2. The Netherland: Delft Universityof Technology and PLAXIS, 2008.
[15] Schanz T, Vermeer PA, and Bonnier PG, "Thehardening-soil model: formulation and verification,"
Beyond 2000 in Computational Geotechnics,
Brinkgreve RBJ, Editor. Rotterdam: Balkema, 1999, pp.
281-290.

121

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Delay Times of Elastic Waves at Discontinuities in Laminated Specimens
Yota Togashi*, Kazuo Tani*, Tetsuji Okada** and Hiroaki Sato**., *Yokohama National University and**Central Research Institute of Electric Power Industry

ABSTRACT: In order to design rock structures, deformation
properties for discontinuous rock mass should be understood. For
the former method of elastic wave velocity measurement, some PS
logging studies showed that S wave velocities of rock masses, Vs,
were found to be dependent on the RQD values. However, theinfluence of the discontinuities of rock masses was not studied yet.
Therefore, a series of elastic wave velocity measurements wereconducted using the laminated specimens of metals and rock cores toinvestigate the influence of the discontinuities. Results demonstrated
that as number of discontinuity n increase, elastic wave velocity (Vp,Vs) become smaller. It is seemed because delay of running time. As
axial compression stress a increase, delay time at discontinuity td
become smaller. It is expected that as a increase discontinuity
become closed.

n=0

n=1

n=3

n=7

100mm

50mm

25mm

12.5mmPhoto1 Laminated specimens (Aluminum)

Keywords: Ultrasonic wave, Elastic wave velocity,
Discontinuity, Rock, Shear modulus

Table1 Physical properties of material for laminatedspecimens
Materials
Aluminium
Nylon
BasaltMudstone

1. INTRODUCTION
In order to design rock structures, deformation properties of
discontinuous rock mass should be understood. For theformer method of elastic wave velocity measurement, some
PS logging studies showed that S wave velocities of rockmasses, Vs, were found to be dependent on the RQD values[1]- [2]. However, the influence of the discontinuities of rockmasses for elastic wave velocity was not studied yet.
Therefore, a series of elastic wave velocity measurements
were conducted using the laminated specimens of metals androck cores to investigate the influence of the discontinuities.

V p(m/s)
6452
2673
5271
2165

t1
td
t2

Element 2

H2

・
・
・
・

EE
EE

H

Element
n+1

Hn+1

2.2 The model for delay times of elastic waves at
discontinuities
Figure1 shows the model for delay times of elastic waves at
discontinuity td. In this model, discontinuities are modeled byboundaries between element &element. There are two
assumptions for this model. The first was that total time ofwave propagation was continuous from top

G d(MPa) E d(MPa)2.642E+04 7.139E+04
1.282E+03 3.607E+03
1.799E+04 4.810E+04
8.652E+02 2.502E+03

0.351
0.41
0.34
0.45

Element 1

H1

2. METHODOLOGY
2.1 Laminated specimens
Photo1 shows laminated specimens. Table1 shows physicalproperties of materials for laminated specimens. These data ismeasured in continuous specimens (n=0 in Photo1).
Laminated specimens have boundaries which arediscontinuous in horizontal direction to loading plane. As
table1 show, laminated specimens are made from Aluminum,
Nylon, Mudstone, Basalt. Height of specimens H =93.0~100mm.

d

V s(m/s)
3091
1057
2616
677

t

td
td
tn+1

Specimen

Figure1 Model for delay times of elastic waveof specimens to that of bottom. The second was that td isconstant values for any discontinuities in each specimen.
From the second assumption, elastic wave velocity in theelement is described as
H
H
H
(1)
V0  1  2    n 1
t1
t2
t n 1
Continuous equation of running time t equation (2) and heightof the specimen H equation (3) are described as

123

n 1

Hi
 ntd
V
i 1
0

t

(2)

Axial load
Rod
Cell
Cap
Pulse

Wave transmission
sourceCell

Pulse generator

Pulse
generator

Specimen

Oscilloscope
Wave acceptor

Oscilloscope

PedestalComputer

Photo2 Conditions for tests

Figure2 Test apparatus

3

0.50

Aluminium, Pwave, a=0.1MPa

Input voltage

n
0
1
3
7

Amplitude A (V)

Amplitude A (V)

2
1
0
-1

n
0
1
3
7

-2
-3

0

100

200

300

400

0.25

n=0
n=7
n=3

0.00
n=1
Input voltage-0.25

Aluminium, P wave, a=0.1MPa

-0.50

500

0

40

80

120

160

200

Time t (s)Time t (s)Figure3 P wave time-lines for aluminum (left: overall view, right: enlarged view )bonding is 0.05~0.21g. The data by these tests are outputtedin oscilloscope by 100MS/s sampling interval. The data used
for analysis are average values by 128 times stacking.

The definition of velocity, equation (3) is described as
n 1

V

H

H i 1

t
t

i



1
1 ntd

V0 H

(3)

2.3 Ultrasonic wave velocity measurements under axial
compression
Figure2 shows the test apparatus. Photo2 shows conditions oftest. As Fig2 shows, ultrasonic wave velocity measurements
under axial pressure (a ≒ 0.1&0.5MPa) are conducted.
Frequencies of wave transmitter are 100kHz (S wave) and
200kHz (P wave). Bonding is done in boundary between
wave transmitter and acceptor and specimens. Amount of

3 TEST RESULTS
Figure3 shows time-lines taken from wave acceptor foraluminum laminated specimens. Allow lines show reading
points of running times. As the values n increases, reading
points become larger in direction of time axis. And also as thevalues n increases, wave amplitude A becomes smaller. It isshown that the wave energy was damped then running times
are delayed at discontinuity because as the values n increases,
the reading point become less-visible.
Figure4 shows the relationships between n and P wave
velocity Vp. Figure5 shows the relationships between n and S

124

3000
td=0std=0s

6000

Pwave velocity VP(m/s)

Pwave velocity VP(m/s)

7000

td=0.5s

5000

td=1s4000
td=2s3000

td=4s

2000

Aluminium0

td=8sa(MPa)
0.1
0.5

1000
0

1

td=16s

2

3

4

5

6

2500

td=0.5s
td=1s2000

td=2s1500

td=4std=8s

1000
Nylon

0

7

td=16sa(MPa)
0.1
0.5

500

0

1

2

Number of discontinuity n

3

4

5

6

6000

2500

5000

Pwave velocity VP(m/s)

Pwave velocity VP(m/s)

td=0std=0.5s

4000

td=1s3000

td=2std=4s2000
Basalttd=8sa(MPa)
0.1
0.5

1000
0

7

Number of discontinuity n01

td=16s

2

3

4

5

6

td=0std=0.5s
td=1s2000

td=2s1500

td=4s1000

td=8std=16s

Mudstone500
07

a(MPa)
0.1
0.5

0

1

Number of discontinuity n

2

3

4

5

6

7

Number of discontinuity n

Figure.4 the relationships between P wave velocity Vp & n
1200
td=0std=0s

3000

Swave velocity VS(m/s)

Swave velocity VS(m/s)

3500
td=0.5s
td=1s2500

td=2s2000

td=4s1500

td=8s
Aluminium

1000
500
0

td=16sa(MPa)
0.1
0.5

900

td=4s

1

2

3

4

5

6

td=8s600

td=16s
Nylon

300

0

0

7

a(MPa)
0.1
0.5

0

1

3

4

5

6

7

1000

2500

td=0s2000

td=0.5s
td=1std=2s1500

td=4s1000

td=8sBasalttd=16sa(MPa)
0.1
0.5

500

0

1

Swave velocity VS(m/s)

3000

Swave velocity VS(m/s)

2

Number of discontinuity n

Number of discontinuity n0td=0.5s
td=1std=2s

2

3

4

5

6

td=2s600

td=32s

Mudstonea(MPa)
0.1
0.5

200

0

1

td=64s2

Figure5 the relationships between S wave velocity Vs & n

125

td=4std=16s3

4

5

Number of discontinuity n

Number of discontinuity n

td=1std=8s400

07

td=0std=0.5s

800

6

7

0.00100.0005
Aluminium

Aluminium

0.0008td/H
0.0003

1/VS(s/m)

1/VP(s/m)

0.00041

0.00020.00010.0000Regression linea(MPa)0.1
0.5

a(MPa)
0.1
0.5

0

1

3

4

5

1

0.00040.00020.00002

td/H

0.0006

6

7

Regression linea(MPa)0.1
0.5

a(MPa)
0.1
0.5

0

1

2

3

4

5

6

7

Number of discontinuity n

Number of discontinuity n

Figure6 Relationships between 1/V and n for aluminum
15
Aluminum
Nylon
BasaltMudstone
Ryolitic tuff (Tanimoto, 1995)

12

Delay time at discontinuity
for Swave tds (s)Delay time at discontinuity
for Pwave tdp (s)

15

9
6
3
0

0.1

a (MPa)

12
9
6
3
0

1

Aluminium
Nylon
BasaltMudstone0.1

a (MPa)

1

Figure7 Relationships between td and a (left: P wave, right: S wave)

wave velocity Vs. In these figures, model curves are also
shown for the td values of 0 to 64s. As the values n increases,
the elastic wave velocities (Vp, Vs) become smaller. It isbecause delay of running time. It would appear that the wave
amplitude was damped by discontinuity then running time
was delayed.
Figure6 shows the relationships between 1/V & n foraluminum. td are calculated from the shapes of different
material’s regression lines.
Figure7 shows relationships between td & a. As a increase,td becomes smaller values. It is expected that as a increase
discontinuity become closed.
In this figure, there are some data by Tanimoto et al (1995) [3].
Their test apparatus is almost same from ours, but theirstudies differ from ours about specimen’s height and placingof discontinuities (in Tanimoto’s studies, placing of
discontinuities are by central.). Almost same tendency
between our data & Tanimoto’s is shown in figure7, so it isexpected that placing of discontinuity is not dominant for thismodel.

4 CONCLUTIONSIn order to clarifying delay of elastic wave velocity at
discontinuity, a series of elastic wave velocity measurements
were conducted using the laminated specimens of metals androck cores.
1) As number of discontinuity n increase, elastic wave
velocities (Vp, Vs) become smaller. It is seemed because delayof running time at discontinuities.
2) As axial compression stress a increase, delay time at
discontinuity td become smaller. It is expected that as aincrease discontinuity become closed.

5 REFERENCES
[1]

[2]

[3]

126

Sato, H., Higashi, S., Shiba, Y., Sato, K., Takahashi, K., Tsuruga, T,.
“Attenuation characteristics of seismic motion based on earthquake
observation records Part3.” Civil Engineering Research Laboratory
Rep.No.N07013, CREIPI, 2007.
Ishii, K., Tani, K., Okada T. and Sato H. “Elastic wave velocity
measurement of aluminum specimen modeling discontinuity in rockmass” Geo-Kanto, JGS, pp.258-259, 2009 (in Japanese).
Tanimoto, C. and Kishida, K. “Fundamental study on seismic wave
propagating property through several kinds of rock joint under uniaxial
stress,” , Journal of JSCE, No. 523/ III- 32, 49-58, 1995.9.

[4]Togashi, Y., Tani, K., Okada, T. & Sato H. “Ultra sonic velocity
measurement for laminated specimens”, Proc. 41st Japan Sym. on RockMechanics, JSCE, 2012 (subbmitted).

127

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Geotechnical Investigation of Valleys using Lightweight Dynamic Cone
Penetration Test for Landslide Risk Assessment
Takashi Hanaoka, Takashi Tsuchida, Shouichi Kawabata, Shouta Nakagawa, Hiroshima University



ABSTRACT: Counter-measures by using the regulations,information technology and the education to the residents areimportant because it take a tremendous time and cost for
implementing physical disaster prevention measures to all
potentially dangerous areas. The purpose of this study is to develop
the site-investigation of natural valley using lightweight cone
penetration test (LWDCPT), in order to implement the risk
assessment of individual valleys with the slope stability analysis
based on the models of valley obtained by the investigation. A series
of investigation was carried out at four valleys in Higashi-Hiroshima
city, which are put on the list of valleys potentially dangerous for
mud and debris flows during heavy rain. LWDCPT was performed
at both sides of valley with each descending to 20 meters withreference to head of valley until gradient of valleys become below
12.5 degrees, in which the necessary time for the site-investigationby two persons was about eight hours per one valley.

valleys in Higashi-Hiroshima City, and selected four valleys
in as same unit. The areas painted in yellow in Fig. 1 show thepotentially dangerous valleys in Saijo Town of Higashi
Hiroshima City. Fig. 2 shows the selected four valleys, whichis in the red circle in Fig. 1, which are included in one
5km×5km regional mesh for the present risk assessment.
Accordingly, the risks in these four valleys are same in thepresent method, and the purpose of this study is to evaluate
the risks of the four valleys individually.

Keywords: Geotechnical investigation, Lightweight cone
penetration test, Individual natural valley, Ground model
1. INTRODUCTION
In Hiroshima Prefecture, 31,987 valleys and slopes are listed
as areas of dangerous for slope failure and debris flow, andthis is the largest number in Japan. Counter-measures byusing the regulations, information technology and theeducation to the residents are important because it take atremendous time and cost for implementing physical disasterprevention measures to all potentially dangerous areas. In
Hiroshima Prefecture, the prefectural land is divided into 350
unit areas of 5 kilometers square, and in each unit area, thecriteria using the rainfall indexes, such as the soil water index
and rainfall intensity, are determined on the basis of pastrecords of landslide disasters due to heavy rain. In this system,a warning for an expected landslide or an evacuation order ismade for each unit area, however, the risk of individual slopesin each area cannot be shown.
To assess the safety of an individual valley or slope, thegeotechnical slope stability analysis is necessary. However,one major problem in natural slopes is that there has been
little information on geotechnical conditions because of theextreme difficulty associated with ground investigation on thesteep slopes covered with vegetation. The purpose of thisstudy is to develop the site-investigation method of naturalvalley using lightweight cone penetration test (LWDCPT), in
order to implement the risk assessment of individual valleys
with the slope stability analysis based on the models of valley
obtained by the investigation.
We focused on the large number of potentially dangerous

Hiroshima University
1 km

Fig. 1. Valleys at risk of landslide of in the town of Saijo.
Saijo:Toyomaru area

4

3

2
1

Saijo:Sukezane area
500 m

Fig. 2. Valleys where the survey were carried out.
2. GEOTECHNICAL INVESTIGATION
Geotechnical investigation was carried out in order to model
the valleys for the seepage analysis and the slope stability
analysis. The necessary items of modeling are slope gradient,
thickness of surface soil and strength parameters. We explain
how to determine of these items.

1292.1 Light weight cone penetration test
LWDCPT has been designed and developed in France since1991 (Langton, D.D., 1999). The schematic view of
LWDCPT is shown in Figure 3. The mass is about 20 kg, andmainly consists of an anvil with a strain gauge bridge, central
acquisition unit (CAU), and a dialogue terminal (DT). The
below from the hammer to the anvil provides energy input,
and a unique microprocessor records the speed of the hammer
and the depth of penetration. The dynamic cone resistance(qd) is calculated from the modified form of Dutch Formula asshown in Equation (1) (Cassan, M., 1988).
1
MV 2
1 2
1
qd 
P
x90
A 1
M

(1)

Where, x90°= penetration due to one below of hammer by
90°cone,A = area of the cone,
M = the mass of striking hammer,
P = the mass of the struck materials, andV = speed of the impact of the hammer.

Photo 1 shows appearance of LWDCPT. The advantages of
LWDCPT are that the detail measurements are possible bygetting a penetration resistance value with each hitting, andthat in virtue of the lightweight and the simple, even for the
non-specialist of geotechnical investigation, the operation iseasy and safe.
2.2 Selection of the testing points in a valley
A geotechnical investigation of the valleys was carried out
under the following policies.
(1)The headwater area was determined by using
topographical data and a field trip. About six geotechnicalinvestigation points were selected in the headwater area, andalong the valley the geotechnical testing points were selected
at intervals of 20 meters, downward.
(2) A test of the valley was carried out at both sides of thevalley, which are 5 located meters away from the center ofvalley. The tests were carried out the headwater down to thevalley, the gradient of which is larger than 12.5 degree.
(3) The sampling was carried out at both sides of foura (riverhead)

Valley 1

hammer
ハンマー

d
f

h
ヘッド
Anvil
データロガー
DialogueTerminal

巻取り式テープ
Scale
ケーブル

中央演算装置CentralAcquisition Unit

e
i

c

b

g
50m

ロッド
Rod

：Penetration test
：Soil sampling

Fig.4. A example of selection of geotechnical investigation
points (Valley 1).
コーン
Cone

Table 1. Number of geotechnical investigation points.

Figure 3. Outline of LWDCPT.

Number of LWCPT

Valley 1
9Valley 2
4

Valley 3
7

Valley 4
7

Number of soil sampling

4

4

4

4

Photo 2. Site of the geotechnical investigation.
(Valley 4)

Photo 1. Appearance of LWDCPT (Valley 1).
130Cone resistance (MPa)
0 2 4 6 8 10 12

1.5

Valley 1
point aleft side2(a) Point a
Cone resistance (MPa)
0 2 4 6 8 10 12

0.5

0.5

1
Valley 1
point fleft side

1.5

2(f) Point f
Cone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12

Valley 1
point bright side

2

0

0

0.5

0.5

1
1.5

(b) Point b

Valley 1
point gleft side

0.5

Valley 1
point cleft side

1
1.5

2

Valley 1
point cright side

0

0.5

0.5

1
1.5

(c) Point c

Valley 1
point hleft side11.5(h) Point hCone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12
0

0

0.5

0.5

0.5

0.5

1.5
2

Valley 1
point dleft side1
1.5

Valley 1
point dright side

Depth (m)0Depth (m)01Valley 1
point hright side2

2

Cone resistance (MPa)
0 2 4 6 8 10 12

Depth (m)Depth (m)02Cone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12

Depth (m)1.5(g) Point g

Depth (m)0.5Depth (m)Depth (m)0Valley 1
point gright side2

Cone resistance (MPa)
0 2 4 6 8 10 12

0

1
1.5

2

Cone resistance (MPa)
0 2 4 6 8 10 12

1

Valley 1
point fright side

1
1.5

2

1
1.5

2

Depth (m)1Cone resistance (MPa)
0 2 4 6 8 10 1200.5

Depth (m)Depth (m)0.5Valley 1
point eright side2

0
Valley 1
point bleft side

1.5

0

1.5

0

Valley 1
point eleft side

1

Cone resistance (MPa)
0 2 4 6 8 10 121Cone resistance (MPa)
0 2 4 6 8 10 12

Depth (m)1(e) Point e

1.5

2

0.5

2

Valley 1
point aright side

Depth (m)1Depth (m)Depth (m)

0.5

0.5

1.5

0

0.5

0

Depth (m)00Depth (m)Cone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12Cone resistance (MPa)
0 2 4 6 8 10 12

Depth (m)investigation points in each valley.
Fig. 4 shows a example of selected of testing points (Valley1). Table 1 shows the number of geotechnical investigation
points of the four valleys. Photo 2 shows the site of thegeotechnical investigation in Valley 1. The testing points
were nine and the soil sampling points were four in Valley 1.

1
1.5

Valley 1
point ileft side

2

2

1
1.5
2

(i) Point i(d) Point dFig. 5. Penetration resistance qd with depth
131

Valley 1
point iright side

Table 2. Thicknesses of surface soil each point (Valley 1)
left sideright side
average
a
0.60
m
1.24
m
0.92
m
b
1.48
m
0.53
m
1.01
m
c
1.10
m
0.65
m
0.88
m
d
0.66
m
0.53
m
0.60
m
e
0.97
m
0.65
m
0.81
m
f
0.26
m
0.58
m
0.42
m
g
1.13
m
0.46
m
0.80
m
h
0.42
m
0.48
m
0.45
m
i
1.22
m
0.99
m
1.11
m
2.3 Investigation of thickness of surface soil
LWDCPT was continued until over 10 MPa value throughthree hitting in a row. Because, the relationship between thepenetration resistance qd of LWDCPT and Nd value of theportable dynamic cone penetration test (PDCPT) was
proposed by Tsuchida et al. (2011), as shown in Equation (2).

qd 

1
Nd
4

Photo 3. Tube samplingwhere, qd = penetration resistance (MPa), γt = unit weight(kN/m3), and z = depth(m).
The cohesion and internal friction angle of the ground were
calculated by Equation (4) and (5), which were proposed by
Tsuchida et al. (2011),(2)

That is, over 10 MPa of qd is considered to be equal to over 40of Nd value, which can be estimated as the basement
rock.Fig.5 shows penetration resistance qd with depth inValley 1. The thickness of surface soil of each point in Valley
1 was estimated as the average value both sides. Table 2
shows thicknesses of surface soil each point in Valley 1.
2.4 Estimation of strength parameters
Using these results, the cohesion and the internal friction
angle of the ground were estimated using the equations
suggested by Tsuchida et al. (2011). In these equations, thedegree of saturation Sr and the penetration resistance of
LWDCPT qd are necessary. In order to obtain degree ofsaturation Sr, the tube sampling was carried out at the point in
Fig. 4 (Valley 1). The appearance of tube sampling is shownin Photo 3. The corrected penetration qd5 was calculated byusing Equation (3), where qd5 is the penetration resistanceunder 5kPa effective overburden stress and the correction
from qd to qd5 are made as follows:

qd 5  qd  0.01  t  z  5(3)

d  29.9  1.61 lnqd 5   0.142Sr(4)

cd  10.6  1.19  lnqd 5   0.041Sr(5)

where, cd = apparent cohesion of drained condition(kN/m2),
d = apparent internal friction angle of drained condition(degree), and Sr = degree of saturation (%).
Table 3 shows the estimation of strength parameters in Valley
1.
3 MODELING OF VALLEYS BY INVESTIGATION
Fig. 6 shows modeling of four valleys by geotechnicalinvestigation. The average values of the thickness of Valley 1,
Valley 2, Valley 3 and Valley 4 were 0.79 m, 0.70 m, 0.57 m,
and 1.31 m, respectively. The average values of the slope
gradients of Valley 1, Valley 2, Valley 3, and Valley 4 were
14.4 degree, 16.3 degree, 16.1 degree, and 18.1 degree,
respectively. In the present risk assessment method, thesefour valleys were included in a mesh for the assessment. By
carrying out geotechnical investigation shown in this study, it
is possible to model the valley and assess the risk by the

Table 3. Estimation of strength parameters (Valley 1).
a

b

c

d

3

wet density ρ ｔ (g/cm )degree of saturation S r (%)

2c d (kN/m )

f

g

h

i

33.61

3

wet unit weight γ ｔ (kN/m )q d5 (MPa)eφ d (degree)

e
1.35
13.21

1.18
0.93
34.9

0.54
0.99
33.7

0.96
0.94
34.6

1.33
0.92
35.1

0.73
0.97
34.2

1.68
0.90
35.5

0.86
0.95
34.4

1.16
0.93
34.9

0.67
0.97
34.0

9.4

8.5

9.2

9.6

8.8

9.8

9.0

9.4

8.7

132

Photo 4. The steep slope.

Photo 5. The gravel under the ground.

stability analysis individually.
As already discussed, in Hiroshima Prefecture, 31,987
valleys and slopes are listed as areas of dangerous for slope
failure and debris flow. Therefore, it is difficult to investigate
all dangerous areas in a short period with the limited budgetof local government. It is preferable in order to become apractical method that the investigation method is as simple
and inexpensive as possible. As one way, it is considered thatto request volunteers, who are not experts but members ofautonomy disaster prevention organization in the area, toinvestigate the valleys. Thus, the investigation method which
everyone can do and personal error not observed ispreferable.
4 CONSIDERATION
The purpose of this study is to develop the procedure of site
investigation of natural valley using lightweight cone
penetration test (LWDCPT), in order to implement the risk
assessment of individual valleys. From the investigations offour valleys, the following points were found;
1) The operation of investigation in this study is very easy,
and it seems that even a non-experts can carry out theinvestigation based on a manual of this method. However,some investigation points, LWDCPT cannot work well,because of many tree roots or gravels under the ground andthe much steepness of slope. In these cases, non-expert may
not cope. And if the penetration resistance appears large
value, sometimes, it is difficult to judge whether there is agravel of a large diameter at this point, or it reaches bedrockas shown in Photo 5.
2) In this investigation, two LWDCPT were used by two
persons (one person use one machine), and other two
persons took undisturbed and disturbed samples at the site.
It took eight hours to investigate all investigation points inValley 1 which has largest investigation points (9 points) in
the four valleys. Therefore, it is possible to investigate eachvalley in a day. However, in case of investigate points areover 10 points, it may be necessary to increase investigatoror to take two days.

5 CONCLUSIONS
The geotechnical investigation using the LWDCPT was
carried out in order to model natural valley or slope for thestability analysis. The advantages of LWDCPT are that thedetail measurements are possible by getting a penetration
resistance value with each hitting, and that in virtue of thelightweight and the simple, even for the non-specialist of
geotechnical investigation, the operation is easy and safe.
Thickness of surface soil and strength parameters were able tobe estimated. Four valleys, which are same risk assessment inpresent method, were able to be modeled individually. It tookeight hours to investigate all testing points in Valley 1 which
has largest number testing points (9 points) in the four valleys.
In case of investigate points under this, it is possible toinvestigate a valley in a day. However, in case that testing
points over more than 10 points, it is necessary to increase
investigator in order to improve efficiency or divide two days.
6 REFERENCES
[1]
[2]
[3]

[4]
[5]

133

Higashi Hiroshima City:
http://www.city.higashihiroshima.hiroshima.jp/site/bousai/hazard.html
Hiroshima Prefecture,
http://www.pref.hiroshima.lg.jp/page/1171592994610/index.html
Langton, D.D., “The PANDA-lightweight penetrometer for soil
investigation and monitoring material compaction,” GroundEngineering, September 1999, pp. 33-37.
Cassan, M., “Les essays in situ en mechanique des sols, realization etinterpretation,” Eyrolles, Paris, 2nd ed. Vol. 1, pp. 146-151.
Tsuchida T, Athapaththu A.M.R.G., Kano S, Suga K, “Estimation of
in-situ shear strength parameters of weathered granitic (Masado) slopes
using lightweight dynamic cone penetrometer,” Soils and Foundations,vol. 51, No. 3. (to be published), 2011.

465
460

c

Thickness 0.9m
Thickness
1.0m
Thickness 0.9m
Gradient
10.8°
Thickness 0.6mGradient 10.8°

d

450

e

445

f

440

g

435

Thickness 0.4m
Gradient 17.6°
Thickness 0.8m
Thickness 0.4m
Gradient 16.6°
Gradient 18.6°

430

Gradient 12.3°
Thickness 0.8m
Gradient 17.6°
surface

h

i

425
420
415

basement

410
0

20

40

60

80

100

Horizontal direction(m)

120

140

160

375

Vertical direction(m)aValley 2 Average layer thickness0.7 m Average gradient16.3°

370
365

b

360

Thickness0.7 m
Gradient15.0°355dThickness0.6 mcThickness0.85 m
Gradient25.7°
surfaceThickness0.65 m
Gradient11.2°

350

basement

345
0.0

10.0

20.0

30.0

40.0

50.0

60.0

Horizontal direction(m)325aVertical direction(m)320Valley 3

315

Average layer thickness0.6 m Average gradient16.1°bThickness0.6 m

c

310
305

d

Thickness0.6 m
Gradient13.0°

300

Thickness0.45 m
Gradient17.0°

e

f

295

g

Thickness0.5 m
Gradient20.0°
surfaceThickness0.75 m

Thickness0.6 m Gradient15.0°
Thickness0.5 m Gradient12.3°
Gradient13.0°

290
285

basement

280
0.0

20.0

40.0

60.0

Horizontal direction(m)

80.0

100.0

120.0

325

a

320

Vertical direction(m)Vertical direction(m)

455

a

b

Valley 1 Average layer thickness 0.8m, Mean gradient 14.4°

Valley 4 Average layer thickness1.3 m Average gradient18.1°

315
310

b
Thickness2.2 m

c

305

Thickness1.1 m
Gradient16.0°

Thickness1.1 m
Gradient15.0°

300
295
290

g

e

Thickness1.1 m
Thickness0.7 m Gradient19.5°
Gradient16.0°dThickness1.5 m
Gradient19.0°

fsurfacebasement

Thickness1.6 m
Gradient17.0°

285
280
0

20

40

60

Horizontal direction(m)

80

100

Fig. 6. Model cross-section of 4 valleys in Higashi-Hiroshima City.

134

120

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Study on Individual Landslide Risk Assessment of Natural Valleys and Slopes
during Heavy Rain Based on Geotechnical Investigation and Analysis
Shouichi Kawabata. Takashi Tsuchida. Takashi Hanaoka. Shouta Nakagawa., Hiroshima University



ABSTRACT: In Hiroshima Prefecture, the prefectural land isdivided into 350 unit areas of 5 kilometers square, and in each unit
area, the criteria using the rainfall indexes, such as the soil water
index and rainfall intensity, are determined on the basis of pastrecords of landslide disasters due to heavy rain. However, in thissystem, the risk of individual valleys or natural slopes in each areacannot be shown. In this study, a series of investigation was carriedout at four valleys in Higashi-Hiroshima city, which are put on thelist of valleys potentially dangerous for mud and debris flows during
heavy rain. For the conditions continuous rain, the infiltration
analysis during the rain, the ground water seepage analysis and theslope stability analysis were carried out based on the ground models
obtained by the site investigation. The risk of slope failure for eachof four valleys was given as the risk assessment point, and wascompared and discussed.

from the past records of the disasters. In this system, the risksof slope failure and evacuation orders are given for an area 5
km×5 km, and the risks to individual slops cannot be shown.
Accordingly, it can be said that the present risk assessment
method is an empirical one, and is not a theoretical orgeotechnical one. In this study, Geotechnical investigationsof four natural valleys in Higashi Hiroshima City were carriedout using a lightweight dynamic cone penetration test and asoil sampling. In order to improve the present risk assessment
method, landslide risk assessments of these valleys for somepredicted rain conditions were made by using a seepage
analysis, a ground water level analysis, and a slope stability
analysis of each valley.

Keywords: Risk Assessment, Landslide, Site Investigation, GroundWater

2. SELECTION OF VALLEYS

1. INTRODUCTION
Hiroshima Prefecture has 31,987 hazardous areas forlandslide disasters, the largest number in Japan in [1]. In
Hiroshima, a regional soil called Masado soil, a weathered
granite, widely covers the surfaces of natural slopes. From ageotechnical point of view, the failures of natural slopes aremainly caused by rises in groundwater levels and the loss ofinbound shear strength of Masado soil due to both continuous
and intensive seasonal rains.
Fig. 1 shows present risk assessment method and purpose ofthis study. The present regional risk assessment method
utilized by the Hiroshima Prefecture government isthoroughly based on the measured rainfall and therainfall-failure relationship in each 5 km×5 km area obtained

In the selection of valleys for the study, we focused on thelarge number of potentially dangerous valleys in
Higashi-Hiroshima City. Fig.2 shows the selected fourvalleys, which are included in one 5 km×5 km regional meshfor the present risk assessment based on the rainfall index.
Accordingly, the risks in these four valleys are the same as in
the traditional method, but the purpose of this study is toevaluate the risks to the four valleys individually.

Saijo:Toyomaru area

4

3

2
1

Present risk assessment method and purpose of this study
Risk Assessment by Rainfall Index (Hiroshima Pref.)Geotechnical mechanismof landslide disasters

Land in prefecture is divided into 5km×5km areas
Statistical analysis on the relation between rainfall
index and Landslide case records in each area

Intense and Continuous
RainfallDetermination of critical rainfall index (CRI) forlandslide and the criteria for alarm and evacuation

・Gradient of slope
・In-situ shear strength ofslope and the reduction
due to the dissipation ofsuction・Rise of ground water level

Use of predicted rainfall and the criteria
however,Too many dangerous slopes
The accuracy of CRI is about 3% for individual slopes
Detail information on
rainfall (predicted)

Geotechnical Information ofindividual slopes

These effects are notpositively used in thepresent risk assessment

Increased accuracy on risk assessment of natural slopes
Decrease of “No incident” rate in the case of evacuation order
Determination of “Risk Level” for individual slopes and accomplishment of
mitigation action

Saijo:Sukezane area
500 m

Fig. 2. Valleys where the survey were carried out in [2].
3. GEOTECHNICAL INVESTIGATION METHOD OF
NATURAL VALLEYS
Table 1 shows geotechnical investigation points. A
geotechnical investigation of the valleys was carried out
under the following policies.
Table 1. Geotechnical investigation points.

Fig. 1. Valleys at risk of landslide of in the town of Saijo.

penetration test
soil sampling

135

valley1 valley2 valley3 valley4
9
4
7
7
4
4
4
4

(1) The headwater area was determined by using
topographical data and a field trip. About six geotechnicalinvestigation points were selected in the headwater area,
and along the valley the geotechnical testing points wereselected at intervals of 20 meters, downward.
(2) A test of the valley was carried out at both sides of thevalley, which are 5 located meters away from the center ofvalley. The tests were carried out the headwater down tothe valley, the gradient of which is larger than 12.5
degree.
(3) The sampling was carried out at both sides of four
investigation points in each valley.

analysis is that used in the expression of Richards in [6]. The
expression of Richards is given by the following equation:

C     z       k  0z t z (1)

where,  =pressure head(cmH2O), C   = specific moisturecapacity, k   = hydraulic conductivity (cm/s). The rainfall
seepage boundary condition in the surface of the ground is

4. RESULTS OF THE GEOTECHNICALINVESTIGATION
The thickness of the surface soil was estimated by analyzing
the penetration resistance of each geotechnical investigation
point. In Fig.3, the penetration resistance qd of lightweight
dynamic cone penetration test with depth, which wasmeasured at point 9 in Valley 1, is shown in [3]. Slope models
were made using the estimated thickness of the surface soil
and the slope gradients. With these results, a modelcross-section of each valley was made, as shown in Fig.4. The
horizontal distance of the cross-section was determined by
the geotechnical investigation point, and the thickness of thesurface soil came out of the average from the estimated valuesof both sides, which were 5 meters away from the center ofthe valley. The average values of the thickness of Valley 1,
Valley 2, Valley 3, and Valley 4 were 0.8 m, 0.7 m, 0.6 m, and1.3 m, respectively. The average values of the slope gradientsof Valley 1, Valley 2, Valley 3, and Valley 4 were 14.4 degree,
16.3 degree, 16.1 degree, and 18.1 degree, respectively.
Furthermore, the void ratio, the density of soil particles, thedegree of saturation, and the unit weight were obtained by thelaboratory tests of soil samples taken in each valley.
Using these results, the cohesion and the internal friction
angle of the ground were estimated using the equations
suggested by Tsuchida et al. (2011).
With these results, cohesions were estimated to be between
8.2 kN/m2 and 11.2 kN/m2, and the internal friction angles
were between 29.6 degree and 36.4 degree.

Photo 1. Site of the geotechnical investigation.
(Valley 3)00.5

Depth (m)1

Afterwards, the unsaturated layer maintains the volumetricwater content  h and descends toward the lower layer. This
unsaturated layer is called the high moisture content belt
(HMCB). The underground water level is formed in the soil
layer when the HMCB reaches the impermeable layer, andthe groundwater level starts to gradually rise from theimpermeable layer upward, as shown in Fig. 5.
To figure out this process, a one-dimensional unsaturated
seepage analysis was carried out at each point of the valleys.
The primitive equation used for the unsaturated seepage
136

1.5
2

Valley 1
Point b2.5
3

Fig. 3. Penetration resistance qd with depth (Valley 1).
465
460

Vertical direction(m)Fig. 5 shows the model of the rainfall seepage in this study.
When it rains continuously and the rainfall begins to infiltrate
into the ground, the volumetric moisture content of the upper
soil layer rises to a certain value  h from the surface layer.

120Valley 1 Average layer thickness 0.8m, Mean gradient 14.4°455

5. SEEPAGE ANALYSIS

Cone resistance (MPa)
2
4
6
8 10

450
445
440
435
430

i

425
420
415
410
0

a

b

c

Thickness 0.9m
dThickness 0.8mThickness 1.0meThickness 0.6mGradient 10.8°Gradient 10.8°fGradient 12.3°
Thickness 0.8mg
Thickness 0.45mGradient 17.6°
surfacehGradient 17.6°
basement
Thickness 0.8m
Thickness 0.45m
Gradient 16.6°
Gradient 18.6°
20

40

60

80

Horizontal direction(m)

100

120

140

Fig. 4. Model cross-section of Valley 1 in
Higashi-Hiroshima City.

High Moisture
Content Belt

An impermeable layer

An impermeable layer

An impermeable layer

An impermeable layer

Fig. 5. Conceptual diagram of rainfall seepage.

160

200.0

100.0

50.0

0.0
0.000

0.100

0.200

0.300

0.400

Fig. 6. Soil moisture characteristic curve used in this study(Yuri, 2010).
Formation of high moisture content

0
10
20
30
40
50
60
70
Valley 1 b
rainfall intensity
5 mm/hr

80
90
100

0.1

0.15

0.2

0.25

0.3

0.35

0.4

0.45

Rainfall intensity 5 mm/hr
Formation of high moisture content
0
10
20
30

rise in groundwater level(3)

0.00hr
2.35hr
4.70hr
7.05hr
9.40hr
11.75hr
14.10hr
16.45hr18.80hr
21.15hr23.50hr25.85hr
28.20hr
30.55hr
32.90hr
35.25hr
37.60hr
39.95hr
42.50hr44.85hr47.25hr

体積含水率θ
The moisture
content by volume

40
50
60
70

A portion of rain water evaporates from the surface of ground,and some is also intercepted by objects such as trees and thegrasses. The rain water that infiltrates into the ground forms ahigh moisture content belt (HMCB) and descends from thesurface to the bottom. When the HMCB reaches the basement
rock, it flows here as underground water, and it flows out tothe surface of the ground when the water table reaches thesurface. An equilibrium relation exists among the amount of
rainfall supplied in a certain period, the amount saved on thesurface, and the volume of water that flows out from thesurface. Here, it is thought that the amounts ofevapotranspiration and that intercepted by trees and grasses
are very small compared with the amount of the rainfall, andthose were neglected. If all the amounts of rainfall infiltrate in
the ground, the following equation can be derived:

0.500

Moisture content by volume θ

6. GROUND WATER LEVEL ANALYSIS

Rinput  Gw  Gwout

weathering granite(Higashi Hiroshima City)

150.0

rise in groundwater level

where, R= amount of rainfall, q= flow velocity in a surface.
Suction was calculated by using the difference approximation
expression of Richards’ equations. The volumetric moisture
content was calculated by applying the obtained suction value
to the soil moisture characteristic curve, in which the initialpressure head and the total head are determined by theexperiments on weathering granite (Masado) carried out by
Yuri (2010); that is, the saturated volumetric moisture content
was 0.433, the residual volumetric moisture content was
0.158, the saturated hydraulic conductivity was 0.006 cm/sand the initial volumetric moisture content was 0.300. Fig. 6
shows the soil moisture characteristic curve used in thisstudy.
Fig. 7 shows an example of the one-dimensional seepage
analysis at Point b in Valley 1 under a rain intensity of 5
mm/hour. As shown in Fig. 7, HMCB is formed from thesurface, and goes downwards. After 14.1 hours, the HMCB
reaches the bottom of the soil layer, and after the saturated
layer is formed at the bottom after 21.15 hr, the groundwaterlevel rises.
The volumetric moisture content in HMCB is closely related
to the rainfall intensity, and the descending speed of HMCB isdecided by the rainfall intensity regardless of the thickness ofthe surface. The seepage analysis also clarified the link
between rainfall intensity in each thickness of the surface andthe time necessary for the formation of groundwater levels inthe soil layers.

Depth(cm)
深度 (cm)(2)

深度 (cm)Depth(cm) R  q  K  
 1
 z


Suction (cmH2O)

given by the following equation:

80

Valley 1 b
rainfall intensity
20 mm/hr

90
100
0.1

0.15

0.2

0.25

0.3

0.35

0.4

0.00hr
1.00hr
2.00hr
3.00hr
4.00hr
5.00hr6.00hr
7.00hr
8.00hr
9.00hr
10.00hr
11.85hr0.45

体積含水率θ
The moisture
content by volume

Rainfall intensity 20 mm/hr
Fig. 7. Change of volumetric moisture content.
The headwater area

Rinputa

Surface soil

b

c

α

s1

β

γ

s2

Basement rock layers3where, Rinput= amount of rainfall, Gw= amount of rainfall
saved in the layer, and Gwout= net amount of water flowing
out due to the groundwater flow.
137

Fig. 8. High moisture content belt descent process, andgroundwater level rise process.

Rinput  r  s1 cos   t(4)

where, r=rainfall intensity(mm/hr), s1= slope length(m).
The amount saved as underground water in the ground of
Section S1 in Δt is:

Gw 





1
1
ne hwt 1  hw1 s1 cos   ne hwt 1s1 cos 
2
2

(5)

1.2

Point bPoint cPoint dPoint ePoint fPoint gPoint hpoint i1
Groundwater level(m)Fig. 8 showed the computing model of the groundwater level
rise.
Water change for each Δt (hr) at Point a of Fig. 8.The amountof rainfall between the ridge and Point a is:

0.8
0.6
0.4

valley1
rainfall intensity 5 mm/hr

0.2
0
0

5

10

15

20

25

30

35

40

45

50

Time(hr)Fig. 9. Groundwater level with the lapse of time after
HMCB reaches the bottom of lay (Valley 1).

where, ne=effective porosity, hwt 1 =groundwater level of Pointa in t (m), and hw1 =groundwater level rise amount of Point

Sliding front

a in Δt (m).
In addition, the amount where underground water flows ofSection S1 in Δt is:

Gwout  k sin   hwt 1 cos   tHigh moisture belt
Sliding front

ht

(6)
H

where, k= hydraulic conductivity.
By substituting Equations (4), (5), and (6) into Equation (7),
we obtain the following:t  tw1h

2 r  s1  k sin   hwt 1  t
 hwt 1s1  s   h βAn impermeable layer

High moisture content belt descent process(7)

where, hwt 1t =groundwater level of Point a in Δt (m), s = saturated moisture content by volume.

Sliding front

Water change for each Δt (hr) in Point b of Fig. 8.As well as
the Equation (7):t  t
w2

h







2 r  s2  k sin   hwt 2  hwt 1  t
 hwt 2  hwt 1  hwt 1t
s2  s   h 

H

(8)

hw

An impermeable layer
βwhere, hwt 2t =groundwater level of Point b in Δt (m),hwt 1 =groundwater level of Point a in t (m), s2= slope length(m) .
Equation (7) is used in Point a of Fig.8. Equation (8) is used in
the other Points of Fig.8.
Fig. 9 shows the changes with the lapse of groundwater level.

Groundwater level rise process
Fig. 10. High moisture content belt descent process, andgroundwater level rise process.
10.0

Groundwater level formationb

Results of the one-dimensional unsaturated seepage analysis,
the slope stability analysis in the high-moisture belt descentprocess, and the groundwater level rise process wereaccomplished based on the physical property values obtainedin the in situ geotechnical investigation, and on the estimatedstrength parameters. Fig. 10 shows the high moisture belt
descent process and the groundwater level rise process. The
slope stability analysis was done as a slide surface with a138

safety factor Fsc

7. SLOPE STABILITY ANALYSIS AND RISKASSESSMENT

d

i

e

1.0

g
f

h

Valley1
rainfall intensity5 mm/hr

0.1
0

10

20

30

40

50

60

70

80

Time(hr)Fig. 11. Changes with the lapse of time of safety factors ofa slope.

465

Valley 1 Average layer rainfall intensity 5 mm/hr,
thickness 0.8m, Mean gradient 14.4°

460

445

39.7hr440435g25.2hr430

hThickness 0.9m
Thickness
1.0m
27.2hrThickness 0.8meThickness 0.6m Gradient 10.8°Gradient 10.8°fGradient 12.3°
Thickness 0.8m
Thickness 0.45mGradient 17.6°

420

surface

Gradient 17.6°
Thickness 0.8m
Thickness 0.45m
Gradient 16.6°
Gradient 18.6°

i

425

c

d

38.2hr450

415

basement

410
0

20

40

60

80
Horizontal direction(m)

100

120

140

160

Valley 1
375

Thickness 0.6m

Vertical direction(m)aValley 2 Average layer rainfall intensity 5mm/hr
thickness 0.7m , Mean gradient 16.3°

370
365

25.1hr360

Thickness 0.7m
Gradient 15.0°

355

d

c
Thickness 0.65m
Gradient 11.2°

350bThickness 0.85m
Gradient 25.7°

surfacebasement

345
0.0

10.0

20.0

30.0

40.0

50.0

60.0

Horizontal direction(m)Valley 2
325

Vertical direction(m)

315
310

43.0hr305

Thickness 0.6m
Gradient 13.0° f

300
295

g

290
285

b

26.0hrdea19.0hrValley 3 Average layer rainfall intensity 5mm/hr,
thickness 0.6m , Mean gradient 16.1°

320

Thickness 0.75m
Thickness 0.6m Gradient 15.0°
Thickness 0.5m Gradient 12.3°
Gradient 13.0°cThickness 0.45m
Gradient 17.0°

Thickness 0.6m
Thickness 0.5m
Gradient 20.0°
surfacebasement

280
0.0

20.0

40.0

60.0

80.0

100.0

120.0

Horizontal direction(m)Valley 3
325

Valley 4
Average layer rainfall intensity 5 mm/hr,
thickness 1.3m , Mean gradient 18.1°

320
315
Vertical direction(m)Vertical direction(m)

455

a

b

310

Thickness 2.2m
Thickness 1.1m
Thickness 0.7m Gradient 19.5°

d300

Thickness 1.1m
Gradient 15°

295
290

g

285
280
0

f

b
c

81.5hr305

a47.2hrGradient 16.0°e

Thickness 1.5m
Gradient 19.0°
Thickness 1.1m
Thickness 1.6m Gradient 16.0°
Gradient 17.0°
20

40

60

surfacebasement
80

100

120

Horizontal direction(m)Valley 4
Fig. 12. Time in which the safety factor of a slope becomes less than one (Rainfall intensity 5 mm/hr).
139

basement rock and a lower side of the high moisture belt
during the HMCB descent process, and as a slide surface withthe basement rock during the groundwater level rise process.
Slope stability analyses were also carried out in different
rainfall conditions: in the various points of the valleys. thosebeing, rainfall intensities of 5 mm/hr, 20 mm/hr, and 50
mm/hr. Fig. 11 shows the results of the slope stability analysis
with a rainfall intensity of 5 mm/hr in Valley 1. In addition,
Fig. 12 shows the required time in which the safety factor of
the slope becomes less than one, in the points in Valley 1
when the rainfall intensity is 5 mm/hr.
Table 2 shows the risk assessments of the four valleys
relatively. In Table 2, the risk assessment points in eachvalley are given by:・Time when a safety factor of a part of the valley becomesless than 1.0.
・Range of the valley where the safety factor becomes lessthan 1.0
And finally, the total risk assessment value was calculated asa sum of the three risk assessment points. It was shown that
Valley 1 is the most dangerous. However, a social importance,
the number of residences downstream is not thought thisstudy. It will be necessary to investigate social importance inthe future.

8. CONCLUSIONIn Hiroshima Prefecture, the prefectural land is divided into
350 unit areas of 5 kilometers square, and in each unit area,
the criteria using the rainfall indexes, such as the soil water
index and rainfall intensity, are determined on the basis ofpast records of landslide disasters due to heavy rain. However,in this system, the risk of individual valleys or natural slopesin each area cannot be shown. In this study, a series of in-situ
geotechnical investigation of four potentially dangerous
valleys inn Higashi-Hiroshima City were carried out. Modelsof the valleys for the geotechnical analysis were constructedby the investigation results. The one-dimensional infiltration
analysis during the rain, the ground water seepage analysis,
and the slope stability analyses when the rainfall intensities
were 5, 20, and 50 mm/hr were continuously carried out.
Based on these analysis, the risk of the four valleys wereshown quantitatively .The risk of slope failure for each of fourvalleys was given as the risk assessment point, and wascompared and discussed. The methods developed in thisstudy are a useful to assess the landslide risk of individualslopes in potentially dangerous areas.
REFERENCES
[1]
[2]
[3]

Table 2. Relative risk assessment of four valleys.
Valley 1 Valley 2 Valley 3 Valley 4
Maximum slope gradient(°)

18.6

25.7

20.0

19.5

Average thickness of the surface soil(m)

0.8

0.7

0.6

1.3

Time that the safety factor of slope
becomes less than one in rainfall
intensity 5mm/hr(hr)Time that the safety factor of slope
becomes less than one in rainfall
intensity 20mm/hr(hr)Time that the safety factor of slope
becomes less than one in rainfall
intensity 50mm/hr(hr)Risk assessment point 1
1) Time that it becomes less than safety
factor 1（On a scale of 1 to 5）
Risk assessment point 22) Range that it becomes less than safety
factor 1（On a scale of 1 to 5）
Total risk assessment
overall risk assessment（On a scale of 1to10）

[4]
[5]

[6]
25.2

25.1

19.0

47.2

5.2

7.7

5.2

12.2

2.3

3.7

2.4

5.7

4.1

3.8

5

1

5

1

3

3

9.1

4.8

8

4

[7]

140

Higashi Hiroshima City:
http://www.city.higashihiroshima.hiroshima.jp/site/bousai/hazard.html
Hiroshima Prefecture,
http://www.pref.hiroshima.lg.jp/page/1171592994610/index.html
Langton, D.D., “The PANDA-lightweight penetrometer for soil
investigation and monitoring material compaction,” GroundEngineering, September 1999, pp. 33-37.
Cassan, M., “Les essays in situ en mechanique des sols, realization etinterpretation,” Eyrolles, Paris, 2nd ed. Vol. 1, pp. 146-151.
Tsuchida T, Athapaththu A.M.R.G., Kano S, Suga K, “Estimation of
in-situ shear strength parameters of weathered granitic (Masado) slopes
using lightweight dynamic cone penetrometer,” Soils and Foundations,vol. 51, No. 3. (to be published), 2011.
Richards,L.A. , “Capillary Conduction of Liquids in Porous Mediums,”
Physics 1, 1931, pp. 318-333.
Yuri H, “Influence of rainfall characteristics and soil properties on soil
moisture change,” master's thesis of graduate school of Hiroshima
university, 2010, pp. 61-89.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Analyses of Piping under Foundation of Weirs in Different Ground Density by
FEM
OKAJIMA Kenji, Department of Environmental Science and Technology, Mie University



ABSTRACT: Creep flow theories that are Bligh’s and Lane’sequation have been used as the safety criteria against piping underfoundation of weirs. These methods were not able to estimate themechanism leading to piping under foundation of weirs. In this study
the effectiveness of our FEM for piping analysis under foundation of
weirs was verified by model experiments.
These model experiments were carried out in five patterns and hadsame creep length by changing the installation position and length of
cut-off wall in high ground density. These critical water heads ofmodel experiments were different from each pattern. It was clear thatcreep flow theories were not able to predict these differences in highground density.
The finite element analysis employs the elasto-plastic constitutive
equations with a non-associated flow rule and strain
hardening-softening. The constitutive equations based on the yieldfunction of Mohr-Coulomb and the plastic potential function of
Drucker-Prager. The finite element is 4-noded iso-parametricelement with one point integration. The explicit dynamic relaxation
method combined with the generalized return-mapping algorithm isapplied. The elasto-plastic constitutive relations including the effectof the shear band are employed. A simplified and generalizedversion of mesh size-dependent softening modulus method (Tanaka
and Kawamoto, 1989) is used in this study.
Our FEM predicted these critical water heads of model experiments.
Maximum shear strain contour line by our finite element analysis
indicated that shear strain concentrated in similar soil mass as
Terzaghi assumed in the seepage failure equation. The analyses ofpiping under foundation of weirs in different ground density
indicated that the critical water head approach to constant with theposition of cut-off wall in loose ground density and the domain forconcentration of maximum shear strain in same weir shape changed
by each ground density.

a series of model experiments that have same creep length,
and then evaluated these practical safety criteria and thevalidity of the elasto-plastic FEM by applying to theexperiments in high relative density sand.
The condition of the foundation of weir is different in each
weir. So we estimated the impact of relative density in thefoundation of the weir on the critical water head by theelasto-plastic FEM.
2. CREEP THEORIESTo prevent piping at the down-stream side of a weir, practicalmanuals indicate that a safe creep length have to be ensured
under the surface of the weir and along the side of the weir.
The creep length to be ensured must be larger than the values
calculated by two methods.
The first method is Bligh's method.

LB  CB H

Where LB is the creep length that is measured along thebottom face of the weir, CB is Bligh's creep ratio which varies
depending on the type of the foundation soil, and H is thewater head. For example the fine sand CB is 15. The criticalhead is H CB when LB  CB H CB .
The second method is Lane's method.

LL  C L H

Keywords: Piping, Creep theory, FEM, Relative density

Where LL is the weighted creep length.

1. INTRODUCTION

LL   lv  kv / k h  lh

Creep flow theories are applied to the design criteria against
piping of foundation of a weir. Bligh’s creep flow theory wasdeveloped as the empirical equation for the design of floatingtype weirs in permeable layer through many experiences in1910 (Bligh)[1]. After suggestion of this theory it was
indicated that vertical sections of the creep length contribute
more to reduce the danger of piping than horizontal sectionsof the length. In the response to this, Lane (1935)[2]
suggested the weighted creep flow theory. These creep flow
theories were based on the assumption that the cause of
piping was erosion along the contact surface between soils
and weir.
The purpose of this study is the reexamination of thesepractical safety criteria against seepage failure. We conducted

(1)

(2)

(3)

Where, lv is the creep length of vertical direction (inclination
angle of more than 45 degrees), lh is the creep length ofhorizontal direction (inclination angle of lower than 45
degrees). kv is the vertical coefficient of permeability and kv
is the horizontal coefficient of permeability. However, kv / k hhas been used 1/3 customarily. CL is Lane's creep ratio which
varies depending on the type of the foundation soils. For
example the fine sand CL is 7.0. H is the water head. The
critical head is H CL , when LL  CL H CL .
3 MODEL EXPERIMENTS
The experimental apparatus was consisted of a glass-walled141

sand box. The size was 1000mm long, 500mm high and200mm wide. The permeable layers in these modelexperiments were made by using clean sand. The sand was
the Toyoura sand with a specific gravity of 2.64, a mean
diameter (D50) of 0.16 mm and a uniformity coefficient of
1.46. The weir was made of rigid acrylic plates. The weir was
fixed to sand box and was sealed by silicon rubber and siliconadhesion bond to prevent water and sand from spilling out.
The sandpaper was pasted on the bottom and side of the weir
to prevent roofing. The cut-off wall was made of aluminum
plate. The sand layers were prepared by pouring dry sandusing hopper into stored water and deleting air during the soil
particle falling. The high density of the sand layers was
obtained: the relative density was about 85% (Figure1).
After setting up the water levels of both upstream anddownstream side equal, the downstream water level was
lowered incrementally (5mm after an hour). The deformationof the sand layer was measured. When piping or boiling
occurred, the water head was defined to attain the criticalwater head.

81.2% to 88.9%. Critical water heads in each pattern weresimilar water heads. In these model experiments the
reproducibility was observed.
These model experiments had same creep length. Bligh’s andLane’s creep flow theories predict a critical water head withpatterns. However, each critical water head was different
from the other pattern. The result indicated that creep flow
theories were not able to predict the critical water head in thehard foundation.
Depth group10-40-M

20-30-M

10mm

20mm

40mm

Position group
5mm 10-40-L

Length of the weir
80mm

10-40-M
10mm

Length: 1000mm

Drainage cistern

Water cistern

Cut-off wall

Permeable
layer thickness

10mm

40mm

40mm

Figure 2 Patterns of model experiments

Width: 200mm

Penetration depth

10-40-R 5mm

*

40mm

Head

50mm

30mm

10mm

Water

50-0
80mm

*

Table1 Results of model experiments
Relative
CriticalAverage of critical
Piping / Boiling
Density water headwater head

Pattern

10-40-M

Discharged water

20-30-M

Figure 1 The layout of the test apparatus
50-0

The data of a series of model experiments are indicated in
Figure2. All patterns had same creep length in which Bligh’s
creep length is 180mm and Lane’s creep length is 123mm.
These data were obtained by conducting 2 or 3 times in eachexperiment. These experiments are divided into 3 groups(Figure1). The first group was named “Depth group” tochange penetration depth of the weir: 10-40-M, 20-30-M and
50-0 (penetration depth - depth of cut-off wall - the positionof cut-off wall). From these experiments we can evaluate theinfluence of the depth of the weir for piping. The second
group was named “Position group” to change the position of acut-off wall: 10-40-L, 10-40-M and 10-40-R. From theseexperiments we can evaluate the influence of the position of
the cut-off wall for piping.
Table1 shows results of model experiments that are relative
density (%), critical water head, the kind of seepage failure(Piping or Boiling) and average of critical water head. Piping
was observed in some patterns of model experiments. The
heaving was observed because sand ground in down-streamside deformed. Relative densities were about 85% from

10-40-L

10-40-R

%
82
86.3
83.6
84.7
86.1
84.5
88.9
85.1
81.4
87.4
81.2
81.5
82.6
83.1
87.4

mm
80
90
90
140
145
125
250
251
271
70
75
75
155
175
180

mm
PipingPipingPipingPipingBoiling
PipingBoiling
Boiling
Boiling
PipingPipingPipingBoiling
Boiling
Boiling

87

137

258

73

170

4 SEEPAGE FAILURE ANALYSES BY
ELASTO-PLASTIC FEM
4.1 Finite element method and analysis conditions
In this study the finite element analysis consisted of two steps.
The first step is the seepage flow analysis by FEM. The
second is the seepage failure analysis by the elasto-plastic
FEM to input effective stress regarding the seepage force as
the external force.
The finite element analysis employs the elasto-plastic
constitutive equations with a non-associated flow rule andstrain hardening-softening. The constitutive equations based142the critical water head of Position group. The critical waterhead gradually increases with the moving to the downstream
end of the weir in Position group which has same creep length.
Results of FEM predicted each critical water head well.
Figure5 indicates the critical water head of Depth group. The
critical water head gradually increases with the increasing thepenetration depth of the weir in Depth group which has same
creep length. Results of FEM predicted each critical waterhead well.
Results of FEM which penetration depth of weir was 10mm
and the cut-off wall was set at upstream side or middle of thelength of weir were computed higher than results of modelexperiments (10-40-M and 10-40-L). The reason isconsidered that the continuum model might not hold true
when number of particle of sand (the average particle side is
0.16mm) was about 60 at the downstream edge of the weir.
Result of FEM which penetration depth of weir was 50mm
was computed lower than results of model experiment (50-0).
The reason was considered that the friction of the side glass
wall influenced the water head of boiling because it wasobserved that larger sand mass moved during the boiling in
the model experiments of 50-0. These discussions indicatedthat our FEM was effective analysis to compute the criticalwater head of the foundation of the weir.

300

Critical water head (mm)on the yield function of Mohr-Coulomb and the plastic
potential function of Drucker-Prager. The finite element is
4-noded iso-parametric element with one point integration.
The explicit dynamic relaxation method combined with thegeneralized return-mapping algorithm is applied. The
elasto-plastic constitutive relations including the effect of theshear band are employed.
A simplified and generalized version of meshsize-dependent softening modulus method (Tanaka andKawamoto, 1989)[3] is used in this study. A material model
for a real granular material (i.e., Toyoura sand) with a highangle of internal friction is used with the features of nonlinear
pre-peak, pressure-sensitivity of the deformation and strengthcharacteristics of sand, non-associated flow characteristics,
post-peak strain softening, and strain-localization into a shear
band with a specific width. The material model will be briefly
described in this section.
In the elasto-plastic finite element analysis, the material
constants of Toyoura sand are as follow: relative density =88%, residual friction angle = 33 degree. The calibration of
the other elasto-plastic parameter of air-dried Toyoura sand in
the elasto-plastic constitutive model was performed using theplane strain compression tests by Tatsuoka et al (1993)[4].
The analysis was performed using a series of finite element
mesh of each model experiment, as shown in Figure3.
Elements around the weir were consisted of 2mm square
mesh. Elements bordering on the weir were boundary
elements and the friction was set to be equal to the friction
between sand and weir in these elements.

(a) All mesh of 50-0 mesh

250
200

Experiments
FEM
Bligh’s creep length
Lane’s creep length

150
100
50
0

10-40-L

10-40-M

10-40-R

Figure 4 Critical water head of Position group(d) 10-40-L mesh
Figure3 FEM mesh300(c) 10-40-M mesh
Critical water head (mm)(b) 20-30-M mesh(e) 10-40-R mesh

4.2 The validation of this FEM
Result of model experiments and FEM in each groups are
discussed about the effectiveness of FEM and the tendency of
seepage failure of foundation of the weir. Figure4 indicates

250
200

Experiments
FEM
Bligh’s creep length
Lane’s creep length

150
100
50
0

10-40-M

20-30-M

Figure 5 Critical water head of Depth group

143

50-0

5.1 Finite element method and analysis conditions
It was considered that the foundation of the weir had various
densities. In this study the effectiveness of our FEM wasverified on the seepage failure of the foundation of the weir.
The effectiveness of our FEM had been verified on theseepage failure of the ground around the sheet pile by
Okajima et al. (2009)[5]. We considered that our FEM wasable to be applied to the seepage failure problem in different
relative density foundation.
We prepared the set of model cases to estimate the impact ofrelative density in the seepage failure of the foundation of theweir. Shapes of weirs were 10-40-L, 10-40-M, 10-40-R,
20-30-L, 20-30-M and 20-30-R. The tendency of seepage
failure in the difference of the position of the cut-off wall andpenetration depth of the weir was checked. Finite element
meshes which were indicated in Figure 6 were applied toanalyses of 20-30-L and 20-30-R. Conditions of relative
density of the foundation were high(85%), middle(50%) andlow(15%) in each weir shape.
5.2 Impact of relative density in seepage failure of weir
Critical water heads of finite element analyses was shown in
Figure 7. Even when relative density was changed, eachdifference of critical water heads of the group of penetration
depth 20mm and cut-off length 30mm (20-30 group) wasmuch the same difference. On critical water heads of thegroup of penetration depth 10mm and cut-off length 40mm
(10-40 group), each difference of critical water heads inhigh(85%) and middle(50%) relative density was much thesame difference but critical water heads in low(15%) relative
density were almost same.
To estimate the seepage failure mechanism of 10-40 group inlow(15%) relative density, Figure8 showed the maximum
shear strain distribution around the weir of 10-40 group inlow(15%) relative density. The contour lines showed ten lines
from 0.1 to 1.0 of maximum shear strain. Our elasto- plastic
FEM has the frictional hardening-softening functions. Whenthe maximum shear strain reached about 0.1, the frictional
function changed from hardening regime to softening regime.
We evaluated that the shear band develops in these elements
at that time. The concentration of maximum shear strain
contour line reached under downstream edge of the weir in
10-40-L and 10-40-M. This indicated that seepage failure
progressed in the ground of downstream in 10-40-L and10-40-M. On the other hand the concentration of maximum
shear strain contour line reached under upstream edge of theweir in 10-40-R. The seepage failure of 10-40-R which had
shallow penetration depth progressed from upstream. 10-40
group of low(15%) relative density had almost same criticalwater head, but the mechanism of seepage failure of each weir
was different. And critical water heads of 10-40-R in
middle(50%) and high(98%) were near to critical water headsof 20-30-R in middle(50%) and high(98%). Critical water

heads of 20-30-L and 20-30-M and critical water heads of
10-40-L and 10-40-M were near each other. It indicated thatthe seepage failure of the foundation of the weir might dependon penetration depth of downstream side of the weir.

(a) 20-30-L mesh
Figure6 FEM mesh(b) 20-30-R mesh200

10-40L

180
160

Critical water head (mm)5 SEEPAGE FAILURE OF THE FOUNDATION IN
DIFERENT RELATIVE DENSITY

10mm
40mm

10-40M

140
120

10-40R100
20-30L

80

20mm
30mm

60
20-30M40
20

20-30R

0
0

20

40
60
80
Relative density (%)

100

Figure7 Relationship of relative density and critical water
head(a) Maximum shear strain distribution at water head 75mm of
10-40-L in low relative density (15%)(a) Maximum shear strain distribution at water head 80mm of
10-40-M in low relative density (15%)

144

(a) Maximum shear strain distribution at water head 80mm of
10-40-R in low relative density (15%)
Figure8 Maximum shear strain distribution of elasto-plastic
FEM
6 RESULT
Creep flow theories were reexamined by model six patterns
experiments and finite element analyses that had same creep
length by changing the installation position and length of
cut-off wall in this study. These critical water heads of modelexperiments were different from each pattern. It was clearthat creep flow theories were not able to predict the criticalwater head.
Our FEM predicted these critical water heads of modelexperiments.
We estimate the impact of relative density in the seepage
failure of the foundation of the weir. Difference of eachcritical water head depending on position of cut-off wall wasless affected by relative density. When the penetration depthof the weir was shallow, however, difference of each criticalwater head depending on position of cut-off wall was few by
the difference of mechanism of seepage failure.
7 REFERENCES
[1]
[2]
[3]

[4]

[5]

Bligh, W. G., “The practical design of irrigation works”, London
Constable, 1910, pp.162-205
Lane, E.W., “Security from Under seepage Masonry Dams on Earth
Foundations”, Trans. ASCE, vol. 100,1935, pp.1234-1351
Tanaka, T. and Kawamoto, O., “Three dimensional finite element
collapse analysis for foundations and slopes using dynamic relaxation”,
Proc. of Numer. Meth. in Geomech., 1988, pp.1213-1218
Tatsuoka, F., Siddiquee, M. S. A, Park, C. S., Sakamoto, M. and Abe,F., “Modeling Stress-Strain Relations of Sand”, Soils and Foundations,vol.33(2), 1993, pp.60-81
Okajima, K, Tanaka, T, Zhang, S and Komatsu, T, “Model experiments
and elasto-plastic finite element analysis about seepage failure of sandbehind fixed sheet pile”, Transactions of the JSIDRE, vol.260, 2009,
pp.107-112

145

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Shear Deformation and Failure of Sandy Slope According to Pore Pressure
Generation due to Rainfall Infiltration
Sasahara K., Kochi University, Sakai N. , NIEDABSTRACT: Monitoring of surface displacement of slope has been
widely adopted as time prediction method of shallow landslide due
to rainfall. Modeling of shear deformation according to rainfall
infiltration is necessary for establishment of the method. In order toexamine the constitutive relation for the model, surface
displacement, pore pressure at the base, volumetric water content(hereafter, V.W.C.) and suction, shear strain in the sandy modelslope are monitored during artificial rainfall. Analysis of themonitored data shows that surface displacement proceeds at small
rate under unsaturated condition first, then increases remarkably
with the rise of pore pressure at the base. Unsaturated component of
surface displacement of steeper slope is larger than that of gentler
slope. It is likely to be due to that shear strain with the increase ofsuction is larger in steeper slope while shear strain according to theincrease of pore pressure is larger in gentler slope.

Keywords: Shallow landslide, Rainfall, Shear strain, Porepressure, Suction
1. INTRODUCTION
The modeling of shear deformation of steep slope due torainfall infiltration is necessary for the establishment of time
prediction method of shallow landslide due to rainfall.
Observations of deformation of the slope under artificial
rainfall or specimen under anisotropic condition just before
the failure had been implemented by some researchers [1]-[3],and produced the empirical law between time anddeformation. The law has been adopted as the time prediction
method based on the monitoring surface displacement [2], [3].

Because this method is empirical one, it does not take thechange of stress in the slope into consideration. So it often
failed to predict the time of failure of the slope under suddenchange of rainfall intensity.
Recently some researches [4], [5] tried to observe slope
deformation by Microelectro-mechanical system (MEMS)
tilt-meter under the change of geometry of the slope such as
cutting works, and other researches [6], [7] observe thedeformation of the slope under artificial rainfall. Although
they made much efforts to measure the slope deformation,
they could not examine the mechanical law of the slope
deformation.
In order to examine the constitutive law for the deformationof the slope under rainfall, the monitoring of V.W.C., suction,
shear strain, pore pressure, and surface displacement aremeasured in the sandy model slope under constant rainfall inthis paper. And some consideration are made in order toderive the relation between V.W.C., suction and shear strain,or that between shear strain and pore pressure in the slope
under rainfall infiltration.
2. METHODOLOGY
2.1 Model slope and monitoring devices
Fig.1 shows the plane and longitudinal section of the modelslope and location of monitoring devices with slope angle of30deg. Model slope is made in a flume of 300cm length,
150cm width, and 50cm height at horizontal section, and600cm length, 150cm width, and 50 cm depth at slope section(Fig.2). The flume has vertical steel blades of 1cm heightlocated every 50cm in the longitudinal direction at the base of
the slope in order to prevent slip between the soil mass andthe base. Model slope is made of granite soil (Fig.3). The soil
is filled and compacted horizontally at every 20cm thickness
due to human stepping on the soil, and is managed to keep

Fig.1 Geometry of model slope and location of monitoring
devices147

Fig.2 Flume and model slope

void ratio of 1.46~1.52. Water content of the soil layer is
3.7~4.4%. The soil layer thickness is 50cm so that soil depth
to gravitational direction is 57.7cm The base and upperboundary of the flume is undrained condition, while the lower
boundary is drained condition.
In the model slope, soil moisture sensors (expressed as ‘VW’
in Fig.1) are installed at the depth of 10, 25, 40cm at 100,
300,500cm from the toe of the slope. They has the accuracy of
0.02 (m3/m3). Tensiometers (expressed as ‘T’ in Fig.1) areinstalled at the depth of 10, 25, 40cm at 300cm from the toe ofthe slope. They have accuracy of 1(kPa). Shear strain gauge(expressed as ‘SS’ in Fig.1) is installed at 300cm from the toeof the slope. It is series of tilt meter at every vertical depth of
9.2cm and tilt meters are connected by bolt and nut. So themeters can incline only to slope inclination direction. Shearstrain increment at some depth Δγ is defined as tan(Δθ), whileΔθ is the inclination increment of the tilt meter (Fig.4). Tilt
meter has non-linearity of 0.2 degree which corresponds tothat of 0.0035 for Δγ. Groundwater level (hereafter G.W.L.)
gauges (expressed as ‘P’ in Fig.1) with the accuracy of
1cmH2O are located at the base of the model at 0, 100, 300,
400, 500cm from the toe of the slope. The measurement data

Percentage finer (%)

100
80
60
40
20
0
0.001 0.01 0.1
1Grain size (mm)10

Fig.3 Grain size distribution of granite soil

Increase of increment of tilt-meter :ΔθShear strain increment Δγ = tan(Δθ)

of the gauge at 100, 300, 500cm are used for the examinationin this paper. Extensometers are installed at the surface of theslope. The location is 100cm, 300cm, 500cm from the toe ofthe slope. It has the non-linearity of 0.1mm
2.2 Experimental conditions
2 types of model slope with different inclination 30deg.
(Fig.1), 40deg. are prepared for the examination of theinfluence of the slope inclination to the relation between shear
deformation and rainfall infiltration. The arrangement of
measurement devices in the slope of 40deg. are just same withthat of 30deg. The difference between the slope of 30deg. andthat of 40deg. are soil layer depth to gravitational direction at
the slope section of the model. Soil layer depth togravitational direction of 30deg. is 57.7cm, while that of
40deg. is 65.3cm. Artificial rainfall with the constantintensity of 50 mm/h is sprayed from the rainfall simulator
above the model slope until the onset of slope failure.
3. RESULTES OF EXPERIMENTS
Fig.5 shows time variation of V.W.C., suction and shear
strain in the slope at 300cm from the toe of the slope, G.W.L.
and surface displacement at 100cm, 300cm, 500cm from thetoe of the model slope with 30deg inclination. Slope failure
occurred at 5150sec. in the model slope of 30deg.. V.W.C.
keeps constant just after the start of the experiment, thenmakes rapid increase earlier with shallower depth, and it
continues to be almost constant after rapid rise. After constantvalue, V.W.C. at 40cm depth increases again at 4300sec. andthat at 25cm also increase again at 5000sec., while that at
10cm keeps almost constant. Suction starts rapid decreasing
earliest at 25cm at 500sec., then 40cm at 1300sec., and keeps
almost constant negative value after that. Suction at 10cm
keeps constant just after the start of experiment then makes
rapid decrease at 3000sec. then makes second decrease at
6000sec.. Shear strain at each depth is almost zero until
3700sec., then increase rapidly. Shear strain at 4.6cm, 13.8cm,
23cm, and 50.6cm proceeds to positive while that at 32.2cm,
41.4cm proceeds to negative. Negative value may be due toreaction of tilt meter at this depth against large movement oftilt meters at upper or lower location. Time variation of
G.W.L. and surface displacement at 100cm is almost samewith those of 300cm, 500cm. This fact suggests that shear
deformation and groundwater rise at any cross-section from100cm to 500cm is almost same. G.W.L. continues slightdecreases from the start of the experiment until 4000sec., andthen makes remarkable increase after almost 4000sec..
Surface displacement also makes no increase until 4000sec.,
then increases rapidly. Surface displacement and shear strain
in the slope makes remarkable increase after around 4000sec
when G.W.L. starts rise. It suggests that shear deformation
proceeds according to increase of pore pressure in the slope of
30deg..
Fig.6 shows time variation of V.W.C., suction and shear

Fig.4 Measurement of shear strain by tilt meter

148

0.4
10cm
25cm

0.3
0.2

40cm

0.1
0

0.3
0.2

10cm
25cm

0.1

40cm

0

6

0.08

0.3

3

Suction (kPa)

-3

50
40
30
20
10
0
-10

10cm
25cm

0

40cm

4.6cm
13.8cm23cm
32.2cm41.4cm50.6cm

0.04
0

-0.04

Shear strain

Suction (kPa)Shear strain

0.4

V.W.C. (cm3/cm3)

V.W.C. (cm3/cm3)

0.5

4.6cm
13.8cm23cm
32.2cm41.4cm50.6cm0.2
0.1

10
G.W.L. (cm)

10
100cm
300cm

5
0

400cm

-5

5

-5

3

15

2

100cm
300cm
500cm

1
0
0

2000 4000 6000

400cm

10

100 cm300 cm500 cm

5
0
0

200 0

400 0

600 0

E lap sed time (sec.)

Elapsed time (sec.)
Fig.5 Time variation of V.W.C., suction, shear strain, porepressure, and surface displacement (30 deg.)

100cm
300cm

0

-10
Surface displacement (cm)

G.W.L. (cm)

40cm

0

15

Surface displacement (cm)

10cm
25cm

Fig.6 Time variation of V.W.C., suction, shear strain, pore pressure,and surface displacement (40 deg.)149

Pore pressure head (cmH2O)

Pore pressure head (cmH2O)

20
10
0
-10
0 1 2 3 4 5Surface displacement (cm)

100cm
300cm
500cm

(a) 30deg.

10
5
0
-5
0
5
10
15
Surface displacement (cm)

100cm
300cm
500cm

(b) 40deg.
Fig.7 Relation between surface displacement and pore pressurestrain in the slope at 300cm from the toe of the slope, G.W.L.
and surface displacement at 100cm, 300cm, 500cm from thetoe of the model slope with 40deg inclination. V.W.C. starts
increase earlier at shallower depth, then keeps almost
constant vale within 0.25~0.3. Suction also starts decreaseearlier at shallower depth, then keeps almost constant value
within 0~10 kPa. Shear strain make little variation until
4000sec., then remarkably increase to positive. G.W.L. keeps
negative and continues slight decrease until 4000sec., thenincrease largely. G.W.L. at 100cm, 500cm is lower than thatat 300cm.Surface displacement also shows almost no
variation until 4000sec., then rapidly increases. Remarkableincrease of shear strain and surface displacement starts
around 4000sec. when G.W.L. starts rise also in the case of
the model slope of 40deg.. It suggests that shear deformationin the slope of 40deg. also greatly influenced by generation ofpore pressure.
4. DISCUSSION
4.1 Surface displacement and pore pressure
Fig.7(a) shows the relation between surface displacement andpore pressure head on the base at 100, 300, 500cm from thetoe of the slope of 30deg.. Pore pressure head at the base isequal to G.W.L.. Pore pressure head decreases and then keeps-7~-3 cmH2O just after the start of the experiment. Surface
displacement increases under unsaturated condition withnegative pore pressure at this stage. After the surface
displacement of 0.5~1.4cm, it increases with the increase ofpore pressure head. Surface displacement increment becomeslarger as pore pressure head increase. So the relation between
surface displacement and pore pressure head can be modified
by hyperbolic curve which is often adopted for stress-strain
relation of the element of soil. The unsaturated component
and all surface displacement are 0.5 and 1.7cm for 100cm, 0.4
and 2.8cm for 300cm, 1.4cm and 2.9cm for 500cm. So theratios of unsaturated component to all surface displacement

are 0.29 for 100cm, 0.14 for 300cm, and 0.48 for 500cm inthe slope of 30deg..
Fig.7(b) shows the relation between surface displacement andpore pressure head on the base at 100, 300, 500cm from thetoe of the slope of 40deg.. Similar to the relation of the slopeof 30deg., pore pressure sudden decreases just after theexperiment and keeps -4~0 cmH2O until surface
displacement of 5~8cm. Surface displacement proceeds
under unsaturated condition in this stage. Although porepressure of 300cm fluctuates and rises up to 2cm in this stage,
it can be thought negative. Fluctuation of pore pressure may
be due to error of measurement. Surface displacement
remarkably increases with increase of pore pressure after that.
Even though pore pressure at the start of this stage of 100,
500cm is negative, this stage can be thought to be undergeneration of positive pore pressure. Negative value of porepressure of -4~0 kPa might mean quasi-saturated condition
near the base of the slope. The unsaturated component and all
surface displacement are 5 and 7.9cm for 100cm, 7.8 and10.9cm for 300cm, 8.2cm and 12.2cm for 500cm. So theratios of unsaturated component to all surface displacement
are 0.63 for 100cm, 0.72 for 300cm, and 0.8 for 500cm in theslope of 40deg.. The ratio of unsaturated component is larger
in the slope of 40deg. than that of 30deg..
4.2 Suction and shear strain in the slope
Fig.8(a) shows the relation between suction and shear strain
at the same depth in the slope of 30deg.. Suction decreases
with small increase of shear strain in the soil layer shallower
than 23cm. Especially shear strain at 4.6cm is almost zeroeven after the increase of suction up to 4 kPa. Suction is still
positive after the decrease at those depths. At the depth
shallower than 23cm, shear strain continues to be almost zerowith the increase of suction up to -4~0 kPa of suction, then it
proceeds under almost constant suction. Shear strain proceeds
to positive at 23, 50.6cm while it proceeds to negative at 32.2,
41.4cm with positive constant suction..

150

Suction (kPa)

0

4.6cm
13.8cm23cm
32.2cm41.4cm50.6cm

4
8
12
-0.04

Suction (kPa)

-4

-10
0
10
20
30
40
50

0
0.04 0.08
Shear strain

4.6cm
13.8cm23cm
32.2cm41.4cm50.6cm0(a) 30deg.

0.1
0.2
Shear strain

0.3(b) 40deg.
Fig.8 Relation between shear strain and suction

Shear strain
Suction (kPa)Shear strain
Suction (kPa)

20

-5

6

Depth (cm)Depth (cm)-30Suction (kPa)03Saturated

40
60
-0.05

0

0.05

0.1

Suction (kPa)
0
5

0
10
20
30
40
50
60

10

Saturated

0

0.1

0.2

Shear strain

Shear strain(a) 30deg.

(b) 40deg.

0.3

Fig.9 Vertical profile of shear strain just before failure

Fig.8(b) shows the relation between suction and shear strain
at the same depth in the slope of 40deg.. Shear strain keeps
almost zero with the increase of suction at first, then proceedswith almost constant suction at all depths. Shear strain
increases under positive suction (unsaturated condition) in thesoil layer shallower than 32.2cm while it increases undernegative suction at 41.4, 50.6cm. Shear deformation proceeds
at unsaturated condition at deeper soil layer in the slope of
40deg. than that of 30deg..
Fig.9(a) shows the vertical profile of shear strain and suction
at 300cm from the toe of the slope of 30deg.. Soil layer withnegative suction is assumed to be saturated. Shear strain at
deeper soil layer is larger than that near surface. Shear strain
at 23~41.4cm depth is negative which might mean inverse
movement of tilt meter at the depth corresponding to large
movement of upper or lower tilt meter. Deeper soil layer issaturated. In the slope of 30deg., shear strain increases largely
at saturated soil layer. Fig.9(b) shows the vertical profile ofshear strain and suction at 300cm from the toe of the slope of
40deg.. Shear strain is larger near surface where suction ispositive, while it is relatively smaller near the base which issaturated. In the slope of 40deg., shear strain increases largely
at unsaturated soil layer.
4.4 Shear deformation under different slope inclination
According to the examination as above, shear deformation atsaturated layer is larger in the gentler slope while shear
deformation at unsaturated layer is larger in steeper slope. In
order to compare the contribution of unsaturated component
to shear deformation at examined section of the slope under
different slope inclination, ratio of unsaturated part of shear151area under different slope inclination is calculated from Fig.9.
Shear area is defined as sum of shear strain from one depth toother depth, and is derived by the equation below.

A



z2

dz

(1)

z1

Here, A: shear area,  : shear strain at some depth, z1, z2:
depth (z1<z2). Shear area under unsaturated layer andsaturated area in each slope is calculated. The ratio of shear
area of unsaturated layer are 0.17 for the slope of 30deg.
while it is 0.58 for 40deg.. So contribution of shear
deformation at unsaturated layer is larger in the steeper slope.
It is same trend with the ratio of unsaturated component of
surface displacement that is larger in the steeper slope.

(3) In the slope, shear strain proceeds under almost constantsuction after rapid increase of suction. Unsaturated soil layer
near surface is thicker in the steeper slope.
(4) Shear strain at saturated zone near bottom is larger thanthat at unsaturated layer in the slope of 30deg., while shear
strain at unsaturated layer near surface is larger in the slope of
40deg..
(5) Contribution of unsaturated shear deformation is larger insteeper slope.

REFERENCES[1][2]

5. CONCLUSION
From the examination as above, the facts as bellows are madeclear.
(1) Surface displacement and shear strain in the slope makes
remarkable increase at the time of rise of G.W.L.
(2) Surface displacement proceeds at unsaturated condition at
first, then increase remarkably with the increase of porepressure at the base. The relation between surface
displacement and pore pressure is hyperbolic just likestress-strain curve of soil element. The ratio of unsaturated
component to all surface displacement is larger in steeper
slope.

[3]
[4]
[5]
[6]
[7]

152

Saitou M and Uezawa H, “Failure of Soil Due to Creep”, Proc. 5thInternational Conference on Soil Mechanics and Foundation
Engineering, vol.1, 1961, pp.315-318.
Saitou M, “Forecasting the Time of Occurrence of a Slope Failure”,
Proc. 6th International Conference on Soil Mechanics and Foundation
Engineering, vol.2, pp.537-541.
Fukuzono T, “A New Method for Predicting the Failure Time of a
Slope”, Proc. IVth International Conference and Field Workshop on
Landslides, 1985, pp.145-150.
Ito K and Toyosawa Y, “Field test of slope failure during slope cuttingwork”, JSCE Journals, vol.65, no.1, 2009, pp.254-265(in Japanese withEnglish abstract).
Uchimura T, et. al., “Simple monitoring method for precaution oflandslides watching tilting and water contents on slopes surface”,
Landslides, DOI 10.1007/s10346-009-0178-z., 2009.
Moriwaki H, et. al., “Failure process in a full-scale landslide
experiment using a rainfall simulator”, Landslides, vol.1, no.4, 2004,
pp.277-288.
Ochiai H, “AFludaized landslide on a natural slope by artificial
rainfall”, Landslides, vol.1, no.3, 2004, pp.211-219.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Load Settlement Relationships of Circular Footings
Considering Dilatancy Characteristics of Sand
Yusuke Tomita, Graduate Student, Dept. of Architecture, Faculty of Environmental and Urban Eng., Kansai Univ., M. Eng.
Tatsuaki Nishigata, Prof., Dept. of Civil, Environmental and Applied System Eng., Faculty of Environmental and Urban Eng., Kansai Univ., Dr. Eng.
Takeshi Masui, Assoc. Prof., Dept. of Architecture, Faculty of Environmental and Urban Eng., Kansai Univ., Dr. Eng.
Shintaro Yao, Prof. Emeritus, Kansai Univ., Dr. Eng.
ABSTRACT: In order to elucidate the mechanical properties ofload-settlement relationships of spread foundations on sand ground,experimental and analytical study on strain hardening and dilatancy
of sand is needed. Here we present the theoretical properties whichwere derived from circular footing model experiment and computer
simulation. The model experiment was carried out through verticalloading on circular footing, and used relative density and tankdimensions as parameters. The quantitative relationship between
load and settlement was analyzed through FEM simulation with
SMP-Cam-Clay model, which is capable of estimating the dilatancy
of sand. The results are summarized as follows; In both cases ofdense and loose sand models, before reaching the ultimate load, theload-settlement relationship obtained from FEM simulation
corresponded with that of obtained from the experiment. Afterreaching the ultimate load, “Terzaghi’s bearing capacity line”
corresponded with the load-settlement relationship obtained fromthe experiment. In the case of medium-dense sand models, somebinding effect of a soil tank rectangle was recognized. The effectwas considered to be exerted by the positive dilatancy of the sand,which occurred steadily with the settlement progresses.

Keywords: Spread Foundation, Sand Ground,Load-Settlement Curve, Constitutive Equation, Dilatancy
1

are given in Table1. The loose sand models were built bypouring sand from containers. The density of the sand models
is a unique-function of the height of free fall of sand.
Medium-dense and dense sand models were built by vibration
provided by an earthquake simulator. The soil tank was madeof steel, which made the tank completely resistant to earthpressure. Teflon sheets with double layers of grease were seton the sides of the soil tank, in order to omit the friction
caused by the experiment ground [7]. To study the effect ofsand dilatancy on mechanical properties of theload-settlement relationship, soil tanks with different
volumes were used in the experiments. The short side of the
big soil tank was measured to be 400 millimeters in Fig.1-1
which shows the experimental device, whereas that of thesmall soil tank was 100 millimeters in Fig.1-2 which showsthe experimental device. The model footing has a circularcross section with diameter of 20 millimeters, and is 80
millimeters in height. Moreover, the model footing is made ofwood, and sandpapers are put on the tip [1]. In the following,

INTRODUCTION

Although the quantitative load-settlement relationships of
spread foundation have been experimentally welldemonstrated on sand ground [1]–[4], the underlying
mechanical properties of the relationships still remained to beelucidated. Especially, since little study has been done on themechanical properties characterized through strain hardening
and dilatancy of sand. Here we present the theoreticalproperties which were derived from circular footing modelexperiment and computer simulation. The model experiment
was carried out through vertical loading on circular footing,and used relative density and tank dimensions as parameters.
The quantitative relationship between load and settlement
was analyzed through FEM simulation with SMP-Cam-Claymodel. In addition, a part of the contents of this paper havebeen reported in [5], [6].
2Table1 Specifications of the Experiment GroundDensity of Soil Particles ρs
2.558g/cm3
Maximum Density ρd max
1.645g/cm3
Minimum Density ρd min
1.337g/cm3

Model Footing
andLoad Cell

Sand Ground

Model Footing
andLoad CellVERTICAL LOADING TEST OF
CIRCULAR FOOTING

2.1 Test Pit and Loading Equipment
The model experiments were performed on Toyoura sand,with different relative densities. The Maximum and minimum
densities of the sand, as determined by the standard procedure,

The Ground Plan：
 5B (Unit：mm)
The Ground Plan：  20B (Unit：mm)
Fig.1-1 Experimental Device (  20B)

153

Fig.1-2 Experimental
Device (  5B)

L-20B
L-5B
M-20B
M-5B
D-20B
D-5B

3

2

0

30

120

D-20B
実験l20B①
D-5B
実験l5B④SD-20B
D-l20B
SD-5B
D-l5B

4
6

Terzaghi’s
Bearing
極限支持力
quCapacity Line

8
10

Fig.4 Comparison of the Load-settlement Relationship obtained from
D-20B and SD-20B, and “Terzaghi’s Bearing Capacity Line”
0.5B(Depth)Element thatstress ratio did
not reach criticalstate line
Element thatstress ratio
reached criticalstate line

(Rectangle)
Fig.5 The Distribution of Stress Ratio obtained from SD-20B(Unit：mm)
0.5B(Depth)3.1.1 In the Case of D-20B[Until p reaches the ultimate load]
The comparison of the load-settlement relationship obtained
from D-20B and SD-20B, and “Terzaghi’s bearing
capacity line” is shown in Fig.4. Here, “Terzaghi’s bearing
capacity line” is calculated as Df S in (1), indicating theline that linked qu as a parameter S. In Fig.4, both p of
SD-20B and D-20B are roughly consistent until S reaches
about 1.0 millimeter. From 1.0 millimeter to 2.0 millimeters,
p of SD-20B is larger than that of D-20B.

qu = αcN c + βγ1 BN γ + γ2 D f N q

90

2

THE INFLUENCE OF SAND DILATANCY ON
P-S RERATIONSHIPS

3.1 In the Case of Dense Sand Models
In the case of dense sand models, sand ground under thefooting seemed to exhibit a continuous behavior with positive
dilatancy until p reaches the ultimate load, whereas the sandbehavior under footing exhibits a discontinuous behavior
with a sign of failure surfaces after p reached the ultimate load.
Therefore, the FEM simulation was performed until p reaches
the ultimate load, and then the rigid-plastic analysis based on
the bearing capacity of Terzaghi’s theory was performed [5].
In the following, SD-20B expresses the analysis
corresponding to D-20B, and SD-5B expresses the analysis
corresponding to D-5B.

荷重p (kN/m )60

0沈下量S (mm)2.3.3 In the Case of Medium-Dense Sand Models
M-20B and M-5B were performed three times, each. As aresult, the p-S relationship showed no significant variation.
The result of a typical case is shown in Fig.3. In any results of
M-20B and M-5B, p steadily increases until S reaches about
1.6 millimeters. Both results are almost consistent until S
reaches 0.8 millimeters. But the stiffness on settlement of
M-5B becomes larger than that of M-20B from 0.8
millimeters to 1.6 millimeters. As a result, p of M-5B is about
1.2 times larger than that of M-20B at S reached about 1.6
millimeters. From 1.6 millimeters to 4.0 millimeters, p
steadily decreases as the settlement progresses where the rateof decrease in p of M-5B is about 3 times larger than that of
M-20B. In any results of M-20B and M-5B, p then steadily
increases again as the settlement progresses, and bothresults appeared to be consistent. In addition, in both resultsof M-20B and M-5B, the ultimate load is confirmed when S
reached about 1.6 millimeters. Such characteristics of the p-S
relationship were similar to that of the p-S relationship which
was derived from D-20B and D-5B. The elevation of sandsurface around the model footing did not appear in any resultsof M-20B and M-5B. Such behavior of sand was similar towhat was derived from L-20B and L-5B.

Element thatstress ratio did
not reach criticalstate line
Element thatstress ratio
reached criticalstate line

(1)

(Rectangle)
Fig.6 The Distribution of Stress Ratio obtained from SD-5B(Unit：mm)

155

to “Terzaghi’s bearing capacity line”. Therefore, sandbehavior is considered to result in a total failure at this stage.
After S reached about 4.5 millimeters, p increases steadilyalong “Terzaghi’s bearing capacity line” as the settlement
progresses, indicating the sand behavior resulted in acontinuous failure.

L-20B

SL-20B

Fig.7(Depth)Comparison of the Load-settlement Relationshipobtained from L-20B and SL-20B
0.5B

Large Deformation
Element
S≧0.26mm

Large Deformation
Element
S≧0.27mm(Rectangle)
Fig.8 The Distribution of “Large Deformation Element”
obtained from SL-20B (Unit：mm)

This is because the progressive failure accompanied by thegeneration of the failure surface occurs. It can be confirmed
from Fig.5 which shows the distribution of stress ratio
obtained from SD-20B when S reached about 2.0 millimeters.
In addition, the element that stress ratio reached critical stateline occurred first when S reached about 1.0 millimeter. In
Fig.5, it is confirmed that the element that stress ratio
reaching critical state line steadily increases from 1.0
millimeter to 2.0 millimeters. Therefore, because thecondition that the boundary of the element changes
discontinuous when stress ratio reached critical state line is
not considered in the case of SD-20B, p of SD-20B is
considered to be larger than that of D-20B from 1.0
millimeter to 2.0 millimeters. From the above, it is consideredthat sand behavior shifts from a continuous behavior based on
SMP-Cam-Clay model to a discontinuous behavior
accompanied by the generation of the failure surface when S
reached about 1.0 millimeter, and that the sheared mass ofsand appears under the footing when S reached about 2.0
millimeters.
[After p reached the ultimate load]
In Fig.4, from 2.0 millimeters to 4.5 millimeters, p decreases
slightly as the settlement progresses. When S reached about
4.5 millimeters, the load-settlement relationship is asymptotic

3.1.2 In the Case of D-5B
[Until p reaches the ultimate load]
The comparison of the load-settlement relationship obtained
from D-5B and SD-5B is shown in Fig.4. Because both p of
SD-5B and D-5B are roughly consistent until S reaches about
1.0 millimeter, the mechanical properties of the p-S
relationship obtained from D-5B is considered to be similar tothat of obtained from D-20B. From 1.0 millimeter to 2.0
millimeters,
both
p of
SD-5B
and
D-5B
are roughly consistent throughout. As a result, the ultimate
load of D-5B is 2.4 times larger than that of D-20B. To
analyze the cause, the comparative studies shown in Fig.5 andFig.6 were conducted. Fig.6 shows the distribution of stressratio obtained from SD-5B. In the case of SD-20B, the area
recognized as sheared mass of sand appears under the footing.
On the other hand, in the case of SD-5B, the area recognized
as sheared mass of sand does not appear. In other words, inthe case of D-5B, positive dilatancy exerts a marked binding
effect of a soil tank rectangle and the dilatancy causes theincrease of mean principal stress as a result. Therefore, from
1.0 millimeter to 2.0 millimeters, the ultimate load of D-5B is
considered to be 2.4 times larger than that of D-20B.
[After p reached the ultimate load]
In Fig.4, from 2.0 millimeter to 4.5 millimeters, p decreases
slightly until p reaches half of the ultimate load. Suchcharacteristic of the p-S relationship is not recognized in thecase of D-20B. Because the failure surface appeared at thesand surface when S reached 4.5 millimeters, the effect ofpositive dilatancy which occurred from 1.0 millimeter to 2.0
millimeters is considered to disappear from 2.0 millimeters to
4.5 millimeters.
3.2 In the Case of Loose Sand Models
In the case of loose sand models, shear failure which occurs
and progresses below edge of the footing is considered tohave a significant influence on the mechanical properties ofthe p-S relationship. Therefore, the element that stress ratio
reached critical state line and shear strain excelled was
defined as “Large deformation element”, and regarded it as agap element in each step, and the FEM simulation was
performed [6]. In the following, SL-20B expresses theanalysis corresponding to L-20B, and SL-5B expresses theanalysis corresponding to L-5B. In addition, because bothresults of SL-20B and SL-5B were almost consistent, only the
results of SL-20B are described as follows.
[Until p reaches the ultimate load]
The comparison of the load-settlement relationship obtained
from L-20B and SL-20B is shown in Fig.7. In Fig.7, both p of
SL-20B and L-20B are almost consistent until S reaches
about 1.0 millimeter. To investigate the stress condition in thesand ground, the distribution of “Large deformation element”
was obtained from SL-20B as shown in Fig.8. In Fig.8,

156

L-20B is shown in Fig.9. In Fig.9, it was obtained theinteresting results that “Terzaghi’s bearing capacity line”
shows a good correspondence with the p-S relationshipobtained from L-20B after S reached 3.0mm millimeters.

L-20B

Ru

3.3 In the Case of Medium-dense Sand Models
A comparison of the ultimate load in the case of eachsand model is shown in Table3. In Table3, Ru is the ratio of
the ultimate load in the case of using  5B against  20B.
Some binding effect of a soil tank rectangle is recognized on
Ru in the case of medium-dense sand model. A binding effectof a soil tank rectangle becomes marked by positive dilatancy
as described in Section 3.1.2. Therefore, in the case of
medium-dense sand model, because the condition ofdilatancy occurrence is located between positive dilatancy
and negative dilatancy, some binding effect of a soil tankrectangle recognized on Ru is considered to be exerted bypositive dilatancy which occurred steadily as the settlement
progresses.

0.96

4

Terzaghi’s Bearing
Capacity Line

Fig.9 Comparison of the Load-settlement Relationship obtained from
L-20B and “Terzaghi’s Bearing Capacity Line”
Table3 Comparison of the Ultimate Load by the Relative Density
Experiment Name

The Ultimate Load(kN/m2)

L-20B

9.51

L-5B

9.17

M-20B

25.01

M-5B

28.70

D-20B

35.66

D-5B

83.91

1.15

2.35

“Large deformation element” appears below edge of thefooting first when S reached 0.26 millimeters, and after S
reached 0.27 millimeters, it appears directly below “Large
deformation element” which occurred when S reached 0.26
millimeters. Because both p of SL-20B and L-20B are almost
consistent until S reaches about 1.0 millimeter, the sandground under the footing is considered to exhibit a behavior
based on SMP-Cam-Clay model until S reaches about 0.26
millimeters. In other words, sand ground under the footing is
considered to exhibit a contractile behavior with negative
dilatancy. From 0.26 millimeters to 1.0 millimeters, it isconsidered that sand ground under the footing exhibits acontractile behavior continuously with negative dilatancy andthat shear failure occurs and progresses at the same time
below edge of the footing.
[After p reached the ultimate load]
In Fig.7, the p-S relationship obtained from SL-20B showed
a good correspondence with that of obtained from L-20B
until S reaches 1.0 millimeter. After S reached 1.0 millimeter,
the simulation of the p-S relationship was carried out based on“Terzaghi’s bearing capacity line,” since SL-20B was
difficult to perform as p of L-20B approached the ultimate
load. Here, the relative density Dr was determined to beconsistent with qu when S reached 3.0 millimeters and theultimate load of L-20B. The relationship between Dr andinternal friction angle φ was estimated from [9]. As a result,
Dr=59%, φ =38.9°, γ1=γ2=1.52g/cm3 were obtained.
These values correspond to the constants representing themechanical properties of the medium-dense sand ground. The
comparison of “Terzaghi’s bearing capacity line” using thesevalues and the load-settlement relationship obtained fromCONCLUSIONSIn this paper, we present the theoretical properties of theload-settlement relationship on sand ground, which were
derived from circular footing model experiment andcomputer simulation. To study the effect of sand dilatancy on
the mechanical properties of the load-settlement relationship,
the model experiment was carried out on relative density andtank dimensions as parameter. The quantitative relationshipbetween load and settlement was analyzed through FEM
simulation with SMP-Cam-Clay model, which is capable ofestimating the dilatancy. The results are summarized asfollows;
1) In the case of dense sand models with large soil tank
(D-20B), it is considered that sand behavior shifts from acontinuous behavior based on SMP-Cam-Clay model toa discontinuous behavior accompanied by the generationof the failure surface when S reached about 1.0
millimeter, and that the sheared mass of sand appears
under the footing when S reached about 2.0 millimeters.
When S reached about 4.5 millimeters, sand behavior isconsidered to result in a total failure. After S reachedabout 4.5 millimeters, sand behavior is considered toresult in a continuous failure.
2) In the case of dense sand models with small soil tank
(D-5B), until S reached about 1.0 millimeter, themechanical properties of the p-S relationship obtained
from D-5B is considered to be similar to that of obtained
from D-20B. From 1.0 millimeter to 2.0 millimeters, it isconsidered that positive dilatancy exerts a marked
binding effect of a soil tank rectangle and the dilatancy
causes the increase of mean principal stress. Accordingly,
the ultimate load of D-5B was 2.4 times larger than thatof D-20B. From 2.0 millimeter to 4.5 millimeters, theeffect of positive dilatancy which occurred from 1.0
millimeter to 2.0 millimeters is considered to disappear.
When S reached about 4.5 millimeters, mechanicalproperties of D-5B are considered to be similar to that of
D-20B.

157

3) In the case of loose sand models with large and small soiltank (L-20B and L-5B), sand ground under the footing is
considered to exhibit a behavior based on
SMP-Cam-Clay model until S reaches about 0.26
millimeters. In other words, sand ground under thefooting is considered to exhibit a contractile behavior
with negative dilatancy. From 0.26 millimeters to 1.0
millimeters, it is considered that sand ground under thefooting exhibits a contractile behavior continuously withnegative dilatancy and that shear failure occurs andprogresses at the same time below edge of the footing.
After S reached 3.0millimters, the results showed that,with the values correspond to the constants representing
the mechanical properties of medium-dense sand ground,“Terzaghi’s bearing capacity line” seemed to correspondwith the p-S relationship obtained from L-20B.
4) In the case of medium-dense sand models (M-20B and
M-5B), some binding effect of a soil tank rectangle wasrecognized on Ru which shows the ratio of the ultimate
load in the case of using  5B against  20B. This isconsidered to be exerted by positive dilatancy which
occurred steadily as the settlement progresses, because
the condition of dilatancy occurrence is located between
positive dilatancy and negative dilatancy.
5

REFERENCES[1] Rei Morimoto et al, “Large-scale plane strain bearing capacity
tests of shallow foundation on sand. (Part 2) ,” 24th Japan
National Conference of Geotechnical Engineering, 1989,
pp.1243-1246[2] K. Terzaghi, “Theoretical Soil Mechanics,” John Wiley ＆Sons. Inc., 1963, pp118-134[3] Vesic, A.S. “Bearing Capacity of Deep Foundations in Sand,”Highway Research Record, Vol.39, 1963, pp112-153[4] E. E. De Beer, “Experimental determination of the shapefactors and the bearing capacity factors of sand, Geotechnique,Vol.20, No.4, 1970, pp387-411[5] Yusuke Tomita et al., “Load Settlement Relationships ofCircular Footings Considering Dilatancy Characteristics ofDense Sand ,” Journal of Structural and Construction
Engineering (Transactions of AIJ), No. 646, 2009,
pp.2263-2270[6] Yusuke Tomita et al., “Load Settlement Relationships ofCircularFootings
Considering
Negative
Dilatancy
Characteristics of Loose Sand ,” Journal of Structural andConstruction Engineering (Transactions of AIJ), No. 661, 2011,
pp.563-570[7] Wei Li et al., “Measurement of soil Displacement around Pile
Tip by Digital Image Analysis,” Geotechnical Engineering in
Urban Construction, Proceeding of the Sino-Japanese
Symposium on Geotechnical Engineering, Beijing, China,
2003, pp.393-400[8] Naotoshi Kashiwa et al. “Displacement Amplitude Dependenceof Effect of Pile Group by Cyclic Lateral Loading Tests on
Large Displacement,” Journal of Structural and Construction
Engineering (Transactions of AIJ), No. 614, 2007, pp.53-60[9] Yorihiko Osaki, “Architectural Foundation Structure,”
gihodobooks, 1991

158

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Thermally Modifying Bentonite for Construction Industry
Abdoullah Namdar, Department of Civil Engineering and Earth Resources, Universiti Malaysia Pahang

Keyword: Soil characteristics; natural soil; ground
improvement; geotechnical problem; soil atomic structure

can result in a slight reduction in bentonite swelling,
probably because of disruption to the cluster-based fabricof bentonite. At high effective stresses the bentonite has
insufficient swelling capacity to force the sand particles
apart, and the sand pore volume thus limits swelling [4].
Bentonite enhanced sand (BES) mixtures are widely usedas barriers to control the movement of liquid from waste
disposal facilities because BES can combine relatively highstrength and low compressibility with very low hydraulicconductivity. This is achieved by using a mixture that
contains sufficient sand to ensure the stability of thecompacted mixture and enough bentonite to seal the voids
between the sand particles. Other benefits are thatcompacted BES containing modest amounts of bentonite isfairly resistant to the effects of desiccation, and thebentonite in BES has a high chemical buffering capacity[5-6].

1. INTRODUCTION
Clay has always played a major role in human life. Clay
raw materials are used and their value recognized in many
economic branches, agriculture, civil engineering andenvironmental studies. This is largely because of their wide
ranging properties, high resistance to atmosphericconditions, geochemical purity, and easy access to theirdeposits near the earth's surface and low price.
Clay minerals, the essential constituents of argillaceous
rocks, can be classified in seven groups according, to theircrystal structure and crystal chemistry. Clay raw materials
are divided in the same way into seven groups. An eighthgroup covers clay ochres and pigments. Further
classification is based on the purpose-made technologicalapplication. It gives information on the application of clay
raw materials or individual clay minerals: in the productionof foods, feedstuffs, beverages, paper, rubber, plastics,
artificial leather, protective coatings for interior andexterior use, pharmaceutics, cosmetics, paints, pencils,
pastels, porcelain, etc [1]. There are many tests have been
performed on bentonite for simulating compaction anddensities. It is heated on one end while hydration with
deionised water takes place by the opposite end to simulate
the thermal gradient created by the heat generated by theradionuclide decay and the hydraulic gradient generated by
the groundwater coming from the host rock into thebentonite barrier [2-3]. It has been considered theengineering behavior of bentonite enhanced sand (BES)
mixtures in relation to their performance as environmental
barriers. Data on the swelling and hydraulic conductivity
are presented. At low effective stresses the bentonite within
BES mixtures swells sufficiently to separate the sandparticles. In such states two factors affect the void ratio
reached by the bentonite after swelling: the ionicconcentration of the pore solution and the bentonite fabricafter compaction. Bentonite swelling is very sensitive to thepore solution concentration because increasing
concentration suppresses the diffuse double layercomponent of swelling. Remoulding during compaction

It has been reported on GMZ bentonite as a potential
material for the construction of engineered barrier in the
Chinese program of geological nuclear waste disposal, for
its high montmorillonite content, high cation exchange
capacity (CEC) and large specific surface etc. Studies onmineralogy and chemical composition, mechanicalproperties, hydraulic behavior, swelling behavior, thermal
conductivity, microstructure and volume change behavior
of GMZ bentonite were performed from 1980s. Based on areview of the former studies, achievements on experimental
and theoretic results obtained on compacted GMZ
bentonite specimens including basic properties, thermal,
hydraulic and mechanical behaviors are presented. Results
show the thermal conductivity of GMZ bentonite and thebentonite based mixtures influenced by its dry density,
water content, mixture of other materials and degree ofsaturation etc. Water retention capacity of highly
compacted GMZ bentonite decreases as the temperature
increases under confined and unconfined conditions. The
hysteretic behavior in the water retention curves of thecompacted GMZ bentonite is not so significant at 20 or
40 °C. The unsaturated hydraulic conductivity ofcompacted GMZ bentonite under unconfined conditions ishigher than that of under confined conditions. This ispossibly induced by the difference in the mechanism of
micro-structural changes during hydration under different
confining conditions. The compaction curves for GMZ
bentonite with different dry densities are clearly step
phased. And the optimum water content for GMZ bentonite
is about 15%. An exponential relationship between
swelling pressure and dry density of highly compacted
GMZ bentonite was determined for the prediction ofswelling pressure. Furthermore, the void ratio after
swelling for unconfined sample also can be predicted using
diffuse double layer (DDL) theory [7]. Conceptually thedesign of high-level radioactive waste (HLW) repositories
in deep geological media includes the construction of anengineered barrier around the waste containers constituted
by a buffer backfill material [8].

ABSTRACT – The application of geology in civil engineering
make possibility of ground improvement. This paper involve withthe thermally modification of construction material based onchanging micro and macro characteristics. The heat could modify
the soil shape, size and chemical composite as well as crystal
structure. In this research work the bentonite subjected to the heatfor 6 hours from 100 ºC to 500 ºC in increment of 100 ºC. The
different techniques and methods have been used for changing soil
micro and macro characteristics. And also the different techniquevalidated this research investigation and the result has been shownthat heat has significant affect on controlling geotechnicalengineering problem. And in the final author would like suggestthat the there is still more scope for continuing this research workon several natural soil and mineral in applying heat for different
time and level for approaching better results.

159

The purpose of the entire this research exercise would be (i)identification under heat bentonite morphology andmechanical properties (ii) and modify bentonite
characteristic for innovation of new material could be use
for mitigation of geotechnical and geo-environment
problems.
2. METHODOLOGY AND EXPERIMENTS
To innovation a new construction material for solving
geotechnical and geo-environment problems aninvestigation on thermally treated bentonite has been
executed. The bentonite has been submitted to the heat for 6hours from 100 ºC to 500 ºC in increment of 100 ºC. The
main research attempt was to modifying construction
material under laboratory condition. The evaluation of bothfor the macro and micro of bentonite characteristics basedon new research work and previous investigation have beentaken systematically trough of laboratory testing. And in
the laboratory triaxial, XRF and SEM tests have been
conducted.
3. RESULTS AND DISCUSSION
The heat has been affected on soil mechanical behavior
based on soil crystal structure and chemical composite andhad significant affect on soil mechanical properties inmacro and micro scale. In this investigation has been
observed that the heat strongly changed soil mechanicalproperties. From the previous investigation in the table 1 isindicated unit weight, optimum moisture content andnatural moisture content of bentonite.

foundation thermally treated bentonite could be acceptablematerial. The soil mechanics experiment result has beenshown that the best safe bearing capacity appeared whensoil is subjected to 500 ºC heat, and the decreasing unit
weight in this level has not affected on final result compare
to when bentonite is under less heat level. In the room
temperature bentonite has 700.06 kN/m2 safe bearing
capacity, when it is submitted to the heat for 500 ºC has
been improved up to 3132.90 kN/m2, it was understood thatthe application of heat on soils could helps in modificationof construction material for improving earth structurestability and impermeability, and also could used in otherindustry. The fig 1 is stress-strain relationship of bentonite
from triaxial test and the fig 2 is [9] also stress-strain
relationship of bentonite from compression tests whensubjected to different level of heat. The results of bothmethods are almost close to each other. When the heat isincreased the stress-strain relationship increased but notlinear and always increasing of heat not resulted in
improving soil bearing capacity in this regard could bringexample of bentonite is submitted to the 400 ºC. This
investigation was for applying heat on the bentonite up to
500 ºC and has also been limited for 6 hours. For the future
investigation could be extend the heat time and level inapplying on bentonite, or other material for approaching
better or optimizing result. For evaluation of changing
bentonite mechanical properties subjected to the heat thechemical element and the morphology of the bentonite
have also been investigated.
2.5

From this research work some other mechanicalproperties of bentonite has been mentioned in the table 2.
The heat is playing in reducing bentonite unit weight andthe interesting issue is the bentonite cohesive characteristic
is increased linearly and from other hand the internal angleof friction is changing nonlinearly. In overall it is find thatthe bentonite with new characteristic is with less weightand more bearing capacity. The changing bentonite crystal
structure due to heat resulted in improved permeability. It
could be suggested that for increasing waterproof of soil

2

Shear stress (kg/cm )300 ºC

1.5
400 ºC

200 ºC1.0

100 ºC
25 ºC

0.5

0.0
0

1

2

3

4

5

6

Strain (%)

7

8

Fig 1 shear stress Vs strain in bentonite clay

2.5

500 ºC
2

Table. 2 the bentonite mechanical properties
Heat
C

Φ [°]
2
3
ºC
)
(KN/m
(KN/m )
RT
12.1
38
0
100
12.2
48
0
200
12.3
56
5
300
12.1
70
10
400
11.6
84
3
500
11.5
98
12

500 ºC

2.0

Shear stress (kg/cm )Table. 1 the bentonite mechanical properties [9]
γ
Heat
OMC
NMC
ºC
(%)
(%)
3
kN/m )
RT
12.1
42.4
13.43
100
12.2
42.5
9.46
200
12.3
42.77
8.9
300
12.1
42.3
5.3
400
11.6
39.2
2.53
500
11.5
38.8
2.33

300 ºC2.0

1.5

200 ºC

400 ºC1.0

100 ºC
25 ºC

0.5

0.0

0

2

4

Strain (%)

68

Fig 2 shear stress Vs strain in bentonite clay [9]1609Table. 3 Chemical element of the bentonite in different
level of heat
Heat
O
Mg
Al
Si
ºC
Wt % 40.71 0.89 8.42 26.97
25
At % 59.18 0.85 7.25 22.33
Wt % 40.91
8.34 26.79
100
At % 59.68
7.22 22.27
Wt % 43.09
8.57 27.26
200
At % 61.27
7.23 22.09
Wt % 36.29
8.05 28.17
300
At % 55.41
7.28 24.50
Wt % 40.06
8.44 27.98
400
At % 58.72
7.33 23.36
Wt % 39.65
9.21 27.67
500
At % 58.29
8.03 23.17
C
1.12
0.73
1.09
0.71
1.14
0.73
0.82
0.56
0.93
0.62
0.50
0.33

Table continued
K
Ti
V
1.90 2.68 0.48
1.13 1.30 0.22
1.95 2.97
1.16 1.45
2.11 2.82
1.22 1.34
1.95 3.03
1.22 1.54
1.73 2.50
1.04 1.22
1.91 2.27
1.15 1.11-

of compaction test the small hydration has been observed. It
is due to developed new bentonite characteristic based onchanging crystal structure. The modification of crystal
structure was responsible for improving bentonite
mechanical properties.

Fe
16.84
7.01
17.95
7.50
15.01
6.12
21.69
9.49
18.36
7.71
18.79
7.91

fig. 3 SEM Photo of bentonite at 25 ºC

The table 3 indicated the chemical composite of thebentonite subjected to the heat from 100 ºC to 500 ºC in
increment of 100 ºC, and the XRF chemical analysis
experiment has been indicated that the Mg and V of thebentonite have been disappeared after applying 100 ºC heat,and the from other hand the remaining elements which are
O, Al, C, K, Ti, Si, Fe showing different level in changing
heat level, and could be understand that the chemicalelements are not responsible for changing soil mechanicalproperties.

Fig. 4 SEM Photo of bentonite processed under 100 ºC
for six hours

The SEM photographs have clearly revealed that thesurface morphology, shape and size of the minerals (figs
3-8). The bentonite under heat for 6 hours from 100 ºC to
500 ºC in increment of 100 ºC selected to study its
morphology modification, the result shown that under all
conditions results are closely similar (fig 3-8) and also same
result is observed about soil chemical composite from the
XRF experiment (table 3) it could be expected that the soil
crystal structure is main reason in modification of soil
mechanical properties. It is observed different color of
bentonite at any level of heat is applied. The changing
bentonite color during increasing heat is due tomodification of crystal structure of bentonite.
It is to be noted that innovation of the better
construction material is possible by application of the heaton bentonite. It is interesting to be mention that in the
bentonite submitted to the heat for 500 ºC after return toroom temperature when mixed with the water to carry out
161

Fig. 5 SEM Photo of bentonite processed under 200 ºC
for six hours

4. CONCLUSION The validation of the results have been proved bycomparing different results
 The bentonite crystal structure was responsible for soil
mechanical properties modification
 It could be suggested this method can apply for
improving construction material characteristic.
 There is different material could be investigated inapplying heat for different time and level for
approaching better results
5. ACKNOWLEDGEMENTS
The author would like to express his thanks to faculty ofcivil engineering, SJCE, Mysore, India, and faculty of earthscience of University of Mysore, India for timely guidance.
Fig. 6 SEM Photo of bentonite processed under
300 ºC for six hours

Fig. 7 SEM Photo of bentonite processed under
400 ºC for six hours

6. REFERENCES[1]. Ji i Konta, Clay and man: clay raw materials in the service of man ,Applied Clay Science, 10, (4), (1995), pp 275-335.
[2]. Martín, M., Cuevas, J., Leguey, S., 2000. Diffusion of soluble salts
under a temperature gradient after the hydration of compacted
bentonite. Applied Clay Science 17, 55–70.
[3]. Villar, M.V., Sánchez, M., Gens, A., 2008b. Behaviour of abentonite barrier in the laboratory: experimental results up to 8 years
and numerical simulation. Physics and Chemistry of the Earth 33(Supplement 1), S476–S485.
[4]. D.I. Stewart, P.G. Studdsb, T.W. Cousens, The factors controllingthe engineering properties of bentonite-enhanced sand, Applied Clay
Science 23 (2003) 97 – 110.
[5]. Tay, Y.Y., Stewart, D.I., Cousens, T.W., 2001. Shrinkage anddesiccation cracking in bentonite – sand landfill liners. Engineering
Geology 60, 263 – 274.
[6]. Yong, R.N., 1999a. Soil suction and soil – water potentials in
swelling clays in engineered clay barriers. Engineering Geology 54,
3 – 13.
[7]. Wei-Min Ye, Advances on the knowledge of the buffer/backfill
properties of heavily-compacted GMZ bentonite, Engineering
Geology 116 (2010) 12-20.
[8]. Villar, M.V., Lloret, A. 2008. Influence of dry density and watercontent on the swelling of a compacted bentonite. Applied Clay
Science 39, 38–49.
[9]. Abdoullah Namdar, et al (2011), Bentonite Thermal Behavior inGeotechnical Engineering, Annals of faculty engineering Hunedoara,
International Journal Of Engineering, Tome IX, Fascicule 1

7. NOMENCLATUREΦ (Degree)= Angle of Friction= Cohesive of Soil
C (KN/m2)
OMC (%)= Optimum Moisture Content
SBC (KN/m2) = Safe Bearing Capacity= Unit Weight (KN/m3)
NMC (%)= Natural Moisture Content

Fig. 8 SEM Photo of bentonite processed under
500 ºC for six hours

162

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Mesh-Free Analysis of Beam on Elastic Foundation
S. M. Binesh, Department of Civil and Environmental Engineering, Shiraz University of Technology, Shiraz, Iran

ABSTRACT: The concept of beam on elastic foundation has been
extensively used by geotechnical engineers for foundation design
and analysis. However most of the studies on the analysis of beamon elastic foundation are devoted to the mesh based methods. In thispaper a mesh-free method is implemented for the analysis of beamon two parameter elastic foundation. The geometry of the problem ismodeled by nodes and the displacement field is constructed by radialbasis functions. The final system of equations is derived by thesubstitution of the displacement field into the weak form of thegoverning equation. The elastic foundation is simulated by theconcept of linkage element and there is no need of node or elementin the traditional sense. At the end of the paper the results of analysis
with the mesh-free method are compared with the results of the finite
element analysis.

mesh definition. There are many mesh-free methods such assmoothed particle hydrodynamics (SPH) [5], element free
Galerkin method (EFGM) [6], reproducing kernel particle
method (RKPM) [7], etc. In this paper the radial basis pointinterpolation method (RPIM) is used for the analysis of beamon elastic foundation.
2.1 Enriched RPIM
According to the enriched RPIM A field function u(x) can beapproximated using both radial and polynomial basis asu  x =nRi  x ai +i

Keywords: Mesh-Free, Beam, Elastic Foundation
1. INTRODUCTION
There are many geotechnical engineering problems that canbe idealized as beams on elastic foundations. This kind ofmodelling helps to understand the soil-structure interaction
phenomenon and predict the contact pressure distribution anddeformation within the medium. The most common theory
for the beam on elastic foundation modelling is the Winkler
approach [1]. However the modelling of soil using the
Winkler's theory was considered inadequate in the handlingof various problems. The main weakness lies in the fact that it
overlooks the shear interaction between the spring elements.
Hence the two-parameter models were proposed somedecades ago [2-3]. The numerical solution for these
two-parameter models are mainly obtained by mesh based
methods such as finite element or finite difference method[4].
In this paper a mesh-free method is implemented for the
analysis of beam on two-parameter elastic foundation. In the
proposed approach the geometry of the beam is modelled bynodes and the displacement field is constructed by radial basis
functions. The two-parameter foundation is simulated by avirtual layer consists of two sets of springs with different
stiffness coefficients. At the end of the paper the results ofanalysis with the mesh-free method are compared with theresults of the finite element analysis.

 P  x b = R  x a + P  x b
j

j

T

T

(1)

j

where R(x) and P(x) are, respectively, the vector of radial andpolynomial basis and n is the number of field nodes in thelocal support domain for point x. Vectors a and b arecoefficients for R(x) and P(x) respectively .
The coefficient vectors a and b are determined by enforcing
Eq. (1) to be satisfied at all the n nodes within the local
support domain. Hence, Eq. (1) can be written as:u  x  = Φ  x  Us

(2)

where Us is a vector of nodal displacements, and Φ  x contains RPIM shape functions for the n local nodes in the
support domain. For details the reader is referred to [8].
3. BEAM ON ELASTIC FOUNDATION MODELING
3.1 Beam ModelingIn the present approach, a beam with any arbitrary thickness
can be simulated readily. As shown in Fig. 1 two or moreparallel sets of nodes may be used to model the beam
structure. Considering the variational (weak) form of the total
potential energy functional for the beam, the discrete form ofequations in mesh-free approach can be written as(3)
K B U B  FB

where

K B   B B T DB B B dΩ

(4)

FB   ΦB T b dΩ   ΦB T T dΓ(5)Ω

2. MESH-FREE METHODS
A new family of numerical methods is developed to get rid of
the deficiencies related to mesh definition. These methods areglobally coined as mesh-free or mesh-less methods and theirmain characteristic is their independency to the traditional

m

Ω

Γ

where Φ and B are respectively, the shape functions and thegradient of shape functions matrices, b is the body force
vector, D is the material matrix,  is the problem domain, T
is the prescribed surface traction and  is the boundary along
which the surface traction is imposed. It should be noted that163

subscript B stands for the beam media.
3.2 Elastic Foundation Modeling
In order to simulate the elastic foundation, the concept oflinkage element is used [9]. As shown in Fig. 2 thetwo-parameter foundation can be considered as a layer with
two stiffness coefficients along two orthogonal directions (i.e.
K S and K n ). According to Fig. 2, the relative deformation
vector  at point P between the top and bottom surfaces canbe related to the displacements of points A and B. Hence it
can be written as

As shown in Fig. 2 the field variable (i.e. displacement) at
point A can be estimated by its nodal values at the nodes
located in the compact support domain of point A. According
to Eqs. (2) , (7) and (8) the displacement of point A can be
written as(9)  B f U SA
where U SA is the displacement vector composed ofdisplacement at all nodes in the compact support domain ofpoint A and(10)
B f  LΦ
Neglecting the normal strain component in the s direction, thestrain vector in the local coordinate can be defined as

1
h

 

(11)

where h is the virtual thickness assumed for the foundation
layer and    ns

Fig. 1 Mesh-free modeling of beam media u  u    S   U A U B   A B  n 
v A  v B (6)

where  S and  n are respectively, the shear and normal
relative displacement at point P. U A and U B are thedisplacement vectors in the local coordinate n-s at points A
and B respectively. However, as the location of point B isfixed, the displacement components of U B (i.e. u B and v B )are both zero. Then equation (6) can be re-written as

 
u 
   S  U A   A 
 n 
v A 

 n  , in which  n is the normal strain inT

the n direction, and  ns is the shear strain. By substituting Eq.
(11) into Eq. (9) the relation between strain and nodal
displacement can be obtained as

1
B f U SAh(12)

To evaluate the stiffness matrix related to the elastic
foundation, the relation between stress and strain in thisregion is also needed. According to Fig. 2 the relation
between stress vector and relative deformation can be writtenas(13)
  Df 
where the stress vector  consists of the normal stress  n(7)and the tangential stress  in the foundation region.

where u A and v A are respectively, the displacement of point    n 

A in the s and n directions.

Matrix D f can also be defined asTK
Df   S
 00 
K n 

(14)(15)

Substituting (11) into (12) gives



1
Df 
h

(16)

Using the variational principle the stiffness matrix of elastic
foundation can be written as

1

1

Κ f =   B f   hDf   B f  dA
h

h

A
Fig. 2 Foundation modeling by linkage element concept(17)

Assuming constant virtual thickness for foundation layer, Eq.
(17) can be written as

Considering the relation between local and global coordinate,
we haveΚ f =  B f Df B f d 

U A  LU A

where  is the length parameter along the foundation layer.
The final matrix form of equations for the whole system ofbeam on elastic foundation can be written as(8)

where L is coordinate transform matrix and U A is thedisplacement vector of point A in global coordinate.

(18)



164

K B  K f U  F(19)Fig. 4 Finite element models of beam on two parameterelastic foundation using 10 and 40 elements

where U and F are nodal displacement and nodal force
respectively.
4. NUMERICAL STUDY

A typical example is investigated to verify the efficiency of
proposed method in this paper. As shown in Fig. 3, a beamwith 10 m length and 1 m thickness is assumed. The elastic
modulus and Poisson ratio of the beam material are 2 GPa and0.25 respectively. The normal and shear stiffness modulus arerespectively, 15000 and 10000 kN/m3. Plane strain conditionis assumed. Two 500 kN concentrated loads are exerted in asymmetric manner at 2 m distance from each end.

Fig. 5 Mesh-free model for the beam on two parameter elastic
foundation

6E-4

Horizontal displacement (m)4E-4

The problem is solved by the finite element program
SIGMAW as well as the proposed mesh-free method. The
finite element modelling consists of two models; 10 elements
model and 40 elements model (Fig. 4). It is obvious that byincreasing the number of elements the accuracy of results will
also increase. Hence this can be a measure to investigate theaccuracy of proposed mesh-free method. As it is depicted inFig. 5 the mesh-free model is constructed by the same numberof nodes as the 10 elements finite element model. The resultsof analyses are shown for the upper and lower surfaces of thebeam in Figs. 6 and 7. Due to the symmetry of the model only
the results for one half of the beam are demonstrated. As it isobvious from the figures, the proposed mesh-free method
offers acceptable results that are even more accurate than the
results of finite element analysis with the same order ofnodes.

0E+0
MFM
FEM (10 ELEMENT)FEM (40 ELEMENT)

-2E-4

-4E-4
0

1

2

3

4

5

Distance from left (m)-5E-3

FEM (10 ELEMENT)

Vertical displacement (m)Fig. 3 Typical problem of beam on two parameter elastic
foundation

2E-4

FEM (40 ELEMENT)
MFM

-6E-3

-7E-3

-8E-3
0

1

2

3

4

5

Distance from left (m)Fig. 6 Horizontal and vertical displacement of the nodes
located at the lower edge of the beam

165

2E-4
FEM (10 ELEMENT)[3]

Horizontal displacement (m)FEM (40 ELEMENT)
MFM

0E+0

[4]
[5]
-2E-4

[6]
-4E-4

[7]
[8]

-6E-4
0

1

2

3

4

5

4

5

[9]

Distance from left (m)-5E-3
FEM (10 ELEMENT)

Vertical Displacement (m)FEM (40 ELEMENT)
MFM

-6E-3

-7E-3

-8E-3
0

1

2

3

Distance from left (m)Fig. 7 Horizontal and vertical displacement of the nodes
located at the upper edge of the beam

5. CONCLUSION

A mesh-free procedure is implemented for the analysis ofbeam on two parameter elastic foundation. The elastic
foundation is simulated by the linkage element concept andthere is no need of node or element in the traditional sense.
The thickness of the beam can be readily adjusted and thenodes can be added or omitted easily. Besides all thesebenefits, the accuracy of results is also acceptable and they
are even better than the finite element results with the sameorder of nodes.

6. REFERENCES[1][2]

Winkler E., in: Die Lehre von der Elastizitat und Festigkeit, Domonicus,
Prague, 1867
Pasternak P.L., "On a new method of analysis of an elastic foundation
by means of two foundations constants", (in Russian), Gasudarstvennoe

166

Izdatelstvo Literaturi po Stroitelstvui Arkhitekture, Moscow, USSR,
1954.
Pronk A.C., "The Pasternak foundation- An attractive alternative for
the Winkler foundation", Proc. of the 5th Int. Conf. on concrete
pavement design and rehabilitation, Purdue Univ. West Lafayette,
Indiana, 1993, Vol. 1.
Bowles J.E., Foundation analysis and design, 5th edition, The
McGraw-Hill Companies, Inc., 1996, Ch. 9.
Libersky L.D., Petschek A.G., "Smoothed particle hydrodynamics withstrength of materials", Proc. of The Next Free Lagrange Conf., 1991,
PP. 248-257.
Belytschko T., Lu Y.Y., Gu L., "Element-free Galerkin methods",
International Journal for Numerical Methods in Engineering, 1994,
Vol.37, PP.229-256.
Liu W.K., Jun S., Zhang Y.F., "Reproducing kernel particle methods",
International Journal for Numerical Methods in Fluids, 1995, Vol.20,
PP.1081-1106.
Liu, G.R., Meshfree Methods: Moving Beyond the Finite Element
Method, CRC Press, Boca, 2003, Ch. 8.
Herrmann L.R., "Finite element analysis of contact problems", ASCE,
Journal of Engineering Mechanics Division, 1978, Vol. 104, No. 5, PP.
1043-1057

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Frictional Performance of Infilled I-Blocks with Geosynthetic Inclusions
Faisal Hj Ali, Department of Civil Engineering, National Defense University of Malaysia
Md. Zahidul Islam Bhuiyan*, Department of Civil Engineering, University of Malaya
Firas A. Salman, Department of Civil Engineering, University of Malaya
Siau Lian Sang, Managing Director, Soil & Slope Sdn. Bhd.


ABSTRACT: Geosynthetic reinforced segmental retaining walls(GR-SRWs) are in a period of development and have achieved
popularity all over the world. Inclusion of geosynthetic layer at theinterface has great influence on interface frictional performance ofsegmental retaining wall units. In this investigation, three (3) typesof geosynthetic reinforcements were chosen and used to find out
their influences on the interface shear capacity of newly designed
and locally produced precast I-blocks. A series of direct shear tests
were conducted using a specially designed and fabricated direct
shear apparatus to assess the frictional characteristics of the blocks
with geosynthetic inclusions. The results presented in this paper
report that flexible geosynthetic reinforcements perform well thanstiff geosynthetics although decrease the interface shear capacity
than no inclusion condition. The presence of geosynthetic layers also
minimizes the localized stress concentrations at the interface.

geogrid (stiff), and a non-woven polyester geotextile(flexible) those which are mostly used in Malaysia for
GR-SRW constructions. Plastic shear pins and natural coarse
aggregate (NCA) were used to increase the interface shear
capacity of the facing systems “to be published” [8]-[9]. A
series of direct shear tests were executed under various
normal loading conditions [3], [10]. Test results were
presented in the form of shear stress-displacement
relationship to compare the effect of different types ofpolymer reinforcements at interface. Shear capacity
envelopes were also drawn using Mohr-Coulomb failure
criteria to outline the angle of friction for different inclusions.
2. MATERIALS

Keywords: geosynthetics; interface shear; segmental block;retaining wall; reinforced soil
1. INTRODUCTION
Segmental concrete units (segmental retaining wall units) as
the facing element for geosynthetic MSE (mechanicallystabilized earth) walls have been frequently used worldwide
for about three (3) decades [1]. They have achieved
popularity due to many advantages in the fields ofgeotechnical engineering. In Malaysia, geotechnicalengineers have been widely exercising geosynthetic
reinforced segmental retaining walls (GR-SRWs) for last
decades [2].
Facing stability is an important issue in the current design
guidelines [3], [4] and it is related to shear and connection
failures. Past research works [5], [6] reported that facing
instability basically occurs due to poor connection strengthand inadequate connection systems.
In GR-SRW constructions, polymer reinforcements are used
to stabilize the backfill soils and facing columns. The
presence of geosynthetic layers at the interface might increaseor decrease the interface shear capacity of segmental concrete
units [1], [7]. It depends on the flexibility of geosynthetic
reinforcements as well as block’s geometry.
In this investigation, three (3) types of geosynthetic
reinforcements were chosen: a knitted polyester (PET)
geogrid (flexible), a high density polyethylene (HDPE)

2.1 Segmental concrete unit
“I” blocks were used as segmental concrete units in thisresearch. “I” blocks are wet cast concrete units (G30), whichhave one center web and the tail/rear flange is extended
beyond the web (Fig. 1). The rear flange is tapered that allows
the blocks to form curve walls. The maximum tapered angleof the “I” block is 11.3 deg. “I” blocks are double open-ended
units and provide a larger hexagonal hollow space inconjunction with two units, and the equivalent holedimensions are about 450 mm in length, 280 mm in width and
300 mm in height. The infill weigh is approximately 93 to 94
kg with the aggregate of bulk density of 1527 kg/m3.The
physical and mechanical properties of the used blocks are
outlined in Table 1.
2.2 Granular infill
The hollow cores between the blocks were infilled with 100%crushed limestone aggregate and lightly compacted. The
maximum and nominal maximum sizes of the aggregate were
25 and 19 mm respectively. The particle size distribution of
the granular infill meets ASTM standard size #57 gradations[11]. The physical properties of infill are given in Table 2.
2.3 Plastic pin
Ultrahigh molecular weight polyethylene (UHMWPE) plastic
bars were used in this investigation as flexible connectors
because of their toughness and flexibility. UHMWPE has also
highest impact strength. The physical and mechanicalproperties of the plastic bars are given in Table 3.

167

2.4 Geosynthetic reinforcements
Three (3) different types of geosynthetic reinforcements wereused in this study because of their high strength and low creep.
Details of the reinforcements are referred as below:
Miragrid GX 80/30 is a knitted uniaxial geogrid preparedfrom high tenacity polyester yarns, and covered with a blackpolymeric coating. The major characteristics are goodconnection capacity with modular blocks and excellent
interface friction behavior, and high tensile strength at low
creep.

Fig. 1 Schematic of used “I” block
Table 1 Physical and mechanical properties of segmental
concrete unit
Property
Dimensions (WxHxL)* in mm
Weight (kg)Oven dry density (g/cm3)%
Water absorption capacity
g/cm3
Moisture content (%)
Net compressive strength (MPa)Value
370x300x500
41-42
2.17
7.1
0.16
3.7
8.0

* W = Width (Toe to heel), H= Height, L= Length (Parallel to the wall face)

Table 2 Physical properties of granular infill
Property
Bulk density (g/cm3)
Specific gravity (SSD)
Void content (%)
Coefficient of gradation, Cc
Fineness Modulus (FM)

Value
1.53
2.63
42
1.15
7.16

E’Grid 90R is an extruded uniaxial geogrid with elongatedapertures. It is made of high density polyethylene (HDPE).
The principal characteristics are good gripping capacity withthe shear connectors of the modular block units and good
creep performance with low strain, and high tensile strength
under constant load.
Rock PEC 75 is a non woven needle punched composite
geotextile consisting of combination between high tenacity
polyester yarns stitched to polypropylene continuous
filaments. It provides high tensile strength at low elongationand high water flow capacity in its plane.
Typical application areas of the geosynthetics (Fig. 2) are
reinforcement of modular block walls, earth walls, slopes andbridge abutments. Table 4 summaries the physical andmechanical properties of the used geosynthetic
reinforcements.

Table 3 Physical and mechanical properties of plastic bar
Property
Yield strength at 23C (MPa)Modulus of elasticity (MPa)Elongation at break (%)
Notched impact strength (kJ/m2)
Density (g/cm3)
Cross section area (mm2)

Value
22
750
>300
No break
0.94
127.66

Fig. 2 Photographs of GX80/30, E’Grid 90R, and RockPEC75 (left to right)

Table 4 Physical and mechanical properties of inclusions
Property

GX80/30
80.0
30.0
34.0
11.0
0.32
23
21

E’Grid 90R
Rock PEC75
90.0
75.0
Short term tensile strength Tc (KN/m)
14.0
MD Tensile strength (KN/m)
45.2
33.8
Strain at MD tensile strength (%)
11.5
10.0
Weight (Kg/m2)
0.55
0.34
MD
240
Aperture size (mm)CD
16
Bond thickness (Tb)
4.1
Thickness (mm)1.40
2.2
Rib thickness (Tr)1.1
Note: MD = machine direction; CD = Cross-machine direction. Unless noted otherwise, data are from manufacturers’
literature.
MDCD
5% strain

168

3. TEST METHODOLOGY
3.1 Experimental device
A specially designed and modified large-scale apparatus wasused to carry out the shear tests of the “I” blocks. A
photograph of the modified test apparatus is illustrated in Fig.
3. It mainly consists of loading frame, hydraulic actuators,
and a fabricated electric hydraulic pump. The verticalactuator was mounted with the loading frame using steel
rollers to allow block movement during the shear test but in
ASTM test protocol the vertical actuator was kept fixing [12].
The vertical and horizontal actuators were capable ofapplying around 45 tons of surcharge load and 130 tons ofpush/pull out force respectively and simultaneously. The
electric hydraulic pump was connected to the actuators withpressure hoses.
A geosynthetic loading clamp was set with horizontal
actuator to apply the tensile load as well as shear load. To
hold the geosynthetic layer at back of the blocks, ageosynthetic gripping clamp was mounted for interface shear
tests. Two (2) pressure transducers were installed over eachhydraulic actuator of 150 mm stroke, and the actuators were
calibrated by using load cell against the pressure transducers.
Two (2) flow regulators were attached with the pump tocontrol the rate of displacement of horizontal (shear) andvertical actuators.
The shear displacements were measured using of two 50 mm
linear variable displacement transducers (LVDTs) with anaccuracy of 0.001mm. Pressure transducers and LVDTs
reading were continuously measured and recorded during thetest by a data logger. The data were recorded at every 10
second interval.

Verticalactuator

3.2 Interface Shear Tests
Two (2) courses of modular block units were used forinterface shear tests. The bottom course consisting of two (2)“I” blocks was installed and braced laterally at the front ofloading frame and the hollow space between the blocks wasfilled up with fresh aggregate, and lightly compacted using asteel rod. Due to tapered rear flange, a small steel plate wasused at the back of the bottom course to hold the compacted
aggregate. One end of the geosynthetic sample was placed
over the bottom course and connected with the plastic shear
pins, and then the other end of geosynthetic was gripped tothe steel clamp for preventing the slippage of thereinforcement layer during the shear tests. A single “I” block
was placed centrally over the running joint formed by the twounderlying units to simulate the staggered construction
procedure used in the field. The double open-ended space ofthe top block was filled up with aggregate and two (2) steel
plates were used to hold the infilled aggregate of the top block.
A photograph of interface shear test arrangement is shown in
Fig. 4.
Normal load was imposed only over the top block throughstiff rubber mat and simulated an equivalent height of thestacked blocks. The shear load was applied against the top
block at a constant rate of 1 mm/min [10]. A steel plate withstiff rubber mat was used with geosynthetic loading clamp toconcentrate the shearing load only over the centrally installed
top block. A horizontal seating load was applied to the top
block to ensure close fitting of the blocks, and after that theload and displacement devices were set to zero. The imposedseating load was 10% of maximum shear strength.
Mohr-Coulomb failure criteria were used to find out interface
shear capacity at ultimate strength criterion.
V = N tan + ɑ.
Where:
V = Interface shear capacity (kPa)N = Normal stress (kPa) = Angle of friction (deg.)ɑ = V interception (apparent cohesion)

Pressure transducers

Horizontal
actuator

Electric pump

Fig. 3 Photograph of test apparatus

Fig. 4 Photograph of interface shear test showing
geotextile sample and gripping system
169

(1)

4. RESULTS AND DISCUSSION
Fig. 5 and 6 illustrate the typical curves for blocks withgeosynthetic inclusions under different normal stresses. From
the Figs. 5 and 6, it is also seen that the ultimate shear stressesof the infilled blocks without any geosynthetic inclusion ishigher than inclusion conditions.
Among the three (3) types of inclusions, polyester geogrid(flexible) perform well than other types of geosynthetics. Thisis happened due to its cushion effects, which minimizes thestress concentration at interface and it also allows theaggregate interlocking through the apertures because of its
grid structures. Figs 5 and 6 demonstrate that the shear stressbehavior of the blocks with HDPE geogrid and polyester
geotextile inclusion is quite same for both normal stresses (54&124 kPa). Even at high normal stress the frictional
performance of the blocks with HDPE geogrid inclusion isalmost equal to those with polyester geotextile (Fig. 6). Dueto the physical structure i.e. thickness and grid structure of
HDPE geogrid, blocks can easily move over each others.
HDPE geogrid works like a friction reducing layer for its
grids, which are stiff and smooth. Its aperture systems also do
not give better interlocking mechanism among the aggregates.
On the other hand, although, the polyester geotextile provide
better cushion at the block’s interface but actually it interrupts
the aggregates interlocking mechanism fully that reduces thefrictional capacity of the blocks with geotextile inclusion.

Fig. 5 Shear stress against displacement
250

Table 7 summarizes performance parameters of the tested “I”
block with and without geosynthetic inclusions. It is seen thatthe flexible geosynthetic inclusions give better angle offriction than stiff geosynthetic layer.

Shear stress (kPa)

150
100
Inclusion (N/A)
Inclusion (flexible PET-geogrid)
Inclusion (stiff HDPE-geogrid)
Inclusion (flexible PET-geotextile)

50
0
0

5

10

15

20

25

Shear displacement (mm)Fig. 6 shows that the ultimate shear stress of the blocks with
polyester geogrid inclusion is almost equal to those without
inclusions. At high normal stress, the shear stress drop of theblocks with HDPE geogrid and polyester geotextile is morethan those with polyester geogrid at low normal stress (Fig.
5).

Fig. 6 Shear stress against displacement
250

200
Shear stress (kPa)Plots of the ultimate interface shear stress against the applied
normal stress are presented in Fig. 7. It is seen that, thepresence of geosynthetic inclusions reduces the ultimate
interface shear capacity of the blocks. Bathurst and Simac [1],and Bathurst et al. [7] observed the same behaviors fordifferent types of blocks with polyester geogrid inclusion.
The ultimate interface shear capacity of the blocks with
polyester geogrid inclusion is closer to no-inclusion condition
because this flexible inclusion improves shear transfer acrossthe block’s interface than others. Fig. 7 also reports that the
reduction in ultimate shear capacity for the inclusions of
HDPE geogrid and polyester geotextile is higher thanpolyester geogrid inclusion. It is influenced by the physicalstructures of the used geosynthetics i.e. flexibility and grid
patterns.

Normal stress = 124 kPa

200

150
100
Inclusion (N/A)
Inclusion (flexible PET-geogrid)
Inclusion (stiff HDPE-geogrid)
Inclusion (flexible PET-geotextile)

50
0
0

50

100

150

200

250

Normal stress (kPa)

Fig. 7 Ultimate shear capacity for “I” blocks
Table 5 Interface shear parameters for different inclusion
conditions
Inclusion
Angle of
Apparentcohesion, ɑ (kPa)friction, (deg.)
N/A
41.2
51.6
Flexible
39.8
46.4
PET-geogrid
Stiff
28.2
53.4
HDPE-geogrid
Flexible
33.2
36.1
PET-geotextile

170

5. CONCLUDING REMARKSThe presence of geosynthetic layer at the facing unit’s(segmental concrete unit) interface reduces the interface shear
capacity. It depends on the flexibility of the used geosynthetic
samples as well as its grid patterns. The angle of friction of
the blocks with polyester geogrid inclusion is higher thanthose with HDPE geogrid and polyester geotextile inclusions.
6. ACKNOWLEDGMENTS
The financial support for this study was provided by a grant(Fundamental Research Grant Scheme - 2010) from the
Ministry of Higher Education of Malaysia awarded to the firstauthor. Materials and technical support were provided by Soil
& Slope Sdn Bhd (Malaysia).
7. REFERENCES[1]

Bathurst RJ, and Simac MR, “Geosynthetic reinforced segmental
retaining wall structures in North America,” in Proc. 5th Int. Conf. on
Geotextiles, Geomembranes and Related Products, 1994, pp. 1-41.
[2] Lee CH, “Design and construction of a 9.6m high segmental wall,” inProc. 2nd Asian geosynthetics conference, 2000.
[3] National Concrete Masonry Association (NCMA) “Design manual forsegmental retaining walls,” Herndon, Virginia, 1997.
[4] Elias V, Christopher BR, and Berg RR, “Mechanically stabilized earth
walls and reinforced soil slopes: Design & construction
guidelines,"FHWA-NHI-00-043, Washington D.C., National Highway
Institute, 2001.
[5] Bathurst RJ, and Simac MR, “Laboratory testing of modular concrete
block - geogrid facing connections,” in Proc. ASTM Symposium on
Geosynthetic Soil Reinforcement Testing, 1993, pp. 32-48.
[6] Soong TY, and Koerner RM, “On the required connection strength ofgeosynthetically reinforced walls,” Geotextiles and Geomembranes,
vol. 15, Aug.-Dec. 1997, pp. 377- 393.
[7] Bathurst RJ, Althoff S, and Linnenbaum P, “Influence of test method
on direct shear behavior of segmental retaining wall units,”
Geotechnical Testing Journal, vol. 31, Mar. 2008, pp. 1-9.
[8] Ali FH, Bhuiyan MZI, and Salman FA, “Effects of mechanicalconnectors on the interface shear capacity of segmental - concrete
blocks in-filled with gravel,” Int. J. of Civil Engineering and Building
Materials, accepted, 2011.
[9] Bhuiyan MZI, Ali FH, and Salman FA, “Frictional behaviour ofsegmental retaining wall units infilled with recycled concrete
aggregate,” in Proc. Symposium on business, engineering andindustrial applications, accepted, 2011.
[10] ASTM D 6916-03, “Standard test method for determining the shear
strength between segmental concrete units,” West Conshohocken, PA,
USA, ASTM International.
[11] ASTM D 448-03a, “Standard classification for sizes of aggregate forroad and bridge construction,” West Conshohocken, PA, USA, ASTM
International.
[12] ASTM D 6916-06c, “Standard test method for determining the shear
strength between segmental concrete units,” West Conshohocken, PA,
USA, ASTM International

171

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Varying Ground Water Level to Minimise Liquefaction Hazards in Urban
Areas
1

Heitor, A.1, Oka, F.2, Kimoto, S.2, Higo, Y.2,
Centre for Geomechanics and Railway Engineering, Faculty of Engineering, University of Wollongong,
2
Dept. of Civil and Earth Resources Eng., Graduate School of Engineering, Kyoto University



ABSTRACT: During a seismic event, the occurrence of liquefaction
causes severe ground deformations that compromise the structural
integrity of many buried elements. Currently, there are a number ofdifferent ground improvement methods that can be adopted to treat
the foundations of the newly constructed structures. However, fewer
remediation methods have been developed for the treatment of theground of already existing superstructures. This study explores theperformance of a cost-effective remediation technique consisting of
inducing ground water level (GWL) depth variation. A 2-D profile
of a soil embankment (i.e. river levee) was selected to model the
GWL variation effect on the deformations associated with the
dynamic loading. The results show that the GWL located at higher
depth, minimizes the overall embankment deformations andcontribute to a significant reduction in the vertical displacements
observed on the top of the embankment.

Keywords: Liquefaction, partially saturated ground,
remediation method, GWL variation,
1. INTRODUCTION
In the event of an earthquake, the occurrence of liquefaction
is often the main cause of severe ground deformations thatcompromise the structural integrity of many buried elementsof urban superstructures, including buildings, foundation
piles, and lifelines. This aspect is of paramount importance,since the vast majority of the largest cities centres in the worldthat are exposed to frequent seismic events, are located by thecoast lines. Furthermore, due to the increasingly urban
development pressure, there is a great deal of superstructures
now constructed in high liquefaction susceptibility terrainsthat were part of earlier reclamation schemes. Currently, there
are a number of different ground improvement methods thatcan be adopted to treat the foundations of the newly
constructed structures, including soil densification, dynamic
compaction or vibroflotation. However, there are few
remediation methods developed for the treatment of theground, on which already existing structures were built. In
fact, a large number of old structures, currently erected inhigh liquefaction susceptibility areas, are without treatment
against liquefaction.
The bubble injection method, proposed in [1], is an example
of a current existing method for liquefaction remediation ofexisting structures. In general, this method consists of
inducing partial saturation in the ground by injecting air
bubbles, which constitutes a very attractive cost-effectiveliquefaction remediation technique. Conversely, in time theair bubbles will dissolve in the ground water and therecurrence interval of the treatment should be adequately

studied. This is the main disadvantage of using this method,since the mechanics of air dissolution in ground water,
necessary to establish the recurrence period are still not wellunderstood.
The method proposed in this study is based on similar concept(i.e. inducing partial saturation in the ground) by artificiallyvarying the ground water level (GWL) (i.e. by pumping
water). Since the GWL is located at greater depth, the groundabove the water level indirectly becomes partially saturated.
And because liquefaction is less likely to occur underpartially saturated conditions [2], [3], the application of thistechnique will likely reduce the overall deformations causedby the occurrence of liquefaction. This method has further
advantages related to the supply of water, since it involves thecontinuous pumping of ground water to maintain the GWL atcertain depth. By doing so, a steady supply of water can be
guaranteed and the water extracted can be either lead tostorage facilities or used for other purposes (i.e. agricultural,industrial use).
The study of the effectiveness of this technique was carriedout numerically, adopting a 2-D profile of a soil embankment(i.e. river levee). Different GWL depths were superimposedin the ground profile and the effect of the GWL change on the
deformations associated with the dynamic loading of 1Hz
sinusoidal wave was investigated. The dynamic numericalanalysis was conducted adopting a finite element code
LIQCA-2D-SF, in which both saturated and partiallysaturated soil theoretical frameworks were included.
2 ELASTO-PLASTIC SIMPLIFIED MODEL FOR
PARTIALLY SATURATED GROUND
The governing equations for gas-fluid-soil coupled problem
can be derived from Biot's type theory of water saturated
porous media based on the continuum mechanics. Herein, thecompressibility of the air is assumed to be very high, or inother words, the three-phase analysis can be simplified into
the soil-water coupled two-phase mixture theory [4], [5]. Anelasto-plastic model with the effect of suction has been
applied to the unsaturated soil using the skeleton stressconcept expressed in (1) as follows,(1) 'ij   ij  S r p f  ij  1  S r  p a ij where Sr is degree of saturation, pf and pa are the pore waterand pore air pressure, respectively and ij is the Kronecker’s
delta.
In the present model the collapse behaviour is described by
the shrinkage of the over consolidation boundary surface dueto the decrease in suction, as shown in (2) and (3), as below:

173

*

 m
0

 mb

(2)
where’m is the mean effective stress,
M*m is the stress ratio*
value at phase transformation,  (0) is the relative stress ratioand ’mb is isotropic consolidation yield stress, which is given
by

 PiC
 
 1  e0 p    1  S I exp   s d  C  1   




P


 

     m bi exp 
 mb

(3)
where ’mbi is initial value of consolidation yield stress ’mb
pre-consolidation pressure, e0 is the initial void ratio,  is thecompression index,  is the swelling index, the vp is theplastic volumetric strain, SI is the is the ratio of reduction ofC
initial suction, sd represents the rate of that change, P isC
initial suction and P is the current suction value. The
governing equations and the elasto-plastic model are
implemented in the finite element code to consider partial
saturation and seepage flow LIQCA-2D-SF [6],[7].
In the numerical analyis the u-p formulation was adopted. For
the discretization of the equations of the motion (orequilibrium of the mixture) FEM (Finite Element Method)
was used, while for discretization of the continuity equationsof the pore fluids (water and gas) FDM (Finite Difference
Method) was used. The time discretization is based on
Newmark's  method, with  and  set at 0.3025 and 0.6,
respectively. The time increment in the calculation was set tobe small enough to guarantee the accuracy of the results
without having large computational time (i.e. 0.01 seconds).
i

between saturated-unsaturated regions the vertical mesh wasrefined, so that the change (herein assumed linear for
simplicity) of degree of saturation and effective stress couldbeadequately
considered.
Thedimensionsofembankment-ground system used in the analysis were set tobe consistent with of a river levee, as in Fig.3. For theboundary conditions it was adopted a horizontal and vertical
fixed condition on the base, equal horizontal displacement
condition on the lateral boundaries and impermeableboundary on base, lateral and top boundaries (Fig.3). The
parameters concerning the soil, the elasto-plastic analysis andthe hydraulic properties are representative of Edosakisa sandand were taken from [4], as shown in Table1.
The results were analysed in terms of top verticaldisplacement and acceleration reduction, as well as in the
Effective Stress Decreasing Ratio (ESDR) minimisation. The
ESDR is a quantitative indicator of the occurrence ofliquefaction and is given as the ratio between the current
effective stress value and the initial mean effective stress, as
follows: ' ESDR  1   m 
  'm 0 (4)in which ’m is the mean effective stress and ’m0 is the initialmean effective stress, and for the unsaturated soil the skeleton
stress is used instead of the effective stress .

3 TWO-DIMENSIONAL NUMERICAL ANALYSIS
To study the influence of the GWL, an embankment-groundsystem was adopted, typically with a total ground depth of
15m. Three cases with GWL’s located at different depths
were adopted as follows:
Case1 ; GWL=0.0m
Case 2 ; GWL=-1.0m
Case 3 ; GWL=-3.0m.
The dynamic input motion used was a 1Hz of frequency
sinusoidal wave with increasing amplitude, following anarithmetic progression scheme, with maximum amplitude of
250 gal (Fig.1). In the liquefaction analysis the definition of
the initial effective is an important step. The profile adopted
was composed of saturated and unsaturated domains, whichmeans that two different approaches have to be considered. In
the saturated domain the effective stress was calculated using
the conventional Terzaghi’s effective stress based on the unit
weight. Conversely, in the unsaturated domain, the skeleton
stress tensor, ’ij, as defined in (1) was adopted. The
horizontal stresses were estimated from the vertical stresses
under the at-rest condition, assuming a coefficient of earthpressure at rest K0 of 0.5 and a Poisson's ratio  of 0.33. In theunsaturated domain, the increase of the negative pore waterpressure (or matric suction) as the depth decrease, wasassumed to follow the relationship given by the soil water
characteristic curve (SWCC) (Fig.2). At the transition
174

Horizontal Acceleration(gal)f b   (*0 )  M m ln

300
200
100
0
-100
-200
-300

Frequency =1.0Hz
0

2

4

6

8

10

12

14

16

18

20

Time (sec)Figure 1: Sinusoidal input motion wave
Figure 2: Unsaturated domain hydraulic model.

Figure 3: Boundary conditions of the embankment-ground system,
ie.GWL=-3m

Top elements-1

Vertical displacement (m)-4.0x10

i

2

3

4

5

6

7

8

GWL= 0.0m depth
GWL= -1.0m depth
GWL= -3.0m depth-1-5.0x10-1-6.0x10

Vertical displacement reduction by 24% 28%-1-7.0x10-1

Figure 4: Top Vertical displacement at t=20sec.
Ground water level -0 m depth
2

Horizontal Acceleration (m/s )

0.4

0.2

0.0

-0.2

-0.4
0

2

4

6

8

10

12

14

16

18

20

Time (sec)Figure 5: Top horizontal acceleration at the top boundary, t=20sec(GWL= 0m, surface).
Ground water level -1 m depth
0.4

0.2

0.0

-0.2

-0.4
0

2

4

6

8

10

12

14

16

18

20

Time (sec)Figure 6: Top horizontal acceleration at the top boundary, t=20sec(GWL= -1m, surface).
Ground water level -3 m depth
2

Horizontal Acceleration (m/s )C

Initial suction: P (kPa)
Suction parameters: SI and sd

0.85
0.00180.006
2.5E-05
19.0
873.28
1.2
0.91
1.12
2200
30
1.1, 1.5
2.04E60.005; 0.001
0.6
1.8 , 3.2
6.6
0.2 ; 0.2

1

-8.0x10

Table 1: Soil and Elasto-plastic parameters
Initial void Ratio: e0
Swelling index: κCompression index: λPermeability: k (m/s)
Density:  (kN/m3)
Normalized Shear modulus: G0/’m
Quasi Over consolidation ratio: OCR*
Phase transformation stress ratio: Mm*
Failure stress ratio: Mf*
Hardening parameter: B0*
Hardening parameter: B1*
Dilatancy parameters: D*, nElastic modulus of water: Kf (kPa)
Plastic, Elastic ref. strain: γrefP＊γrefE＊
Initial degree of Saturation: Sr
van Genuchten parameters:  and n’

02

The results show that the GWL located at higher depth,
minimizes the overall embankment deformations andcontribute to a significant reduction in the verticaldisplacements observed on the top of the embankment (Fig.
4). In fact, the GWL located at a larger depth yielded smaller
vertical displacements on top of the embankment levee.
However, the vertical displacements are reduced by a moresignificant amount for the case with GWL located closer tothe surface at -1m depth (i.e. 24% reduction). In the casewhere GWL was set a deeper level (GWL=-3m) the reductionof the vertical displacements was marginal comparatively tothe case with GWL -1m depth (increase of 4%). These results
indicate that even small GWL depth changes (i.e. up to -1m
depth) are likely to minimise significantly the verticaldisplacement of the superstructures located on highliquefaction susceptibility areas. In Figs. 5, 6 and 7 arerepresented the horizontal acceleration time profiles for the
top right corner of the embankment levee together with theirrespective envelopes. In Fig. 7 the envelopes representativeof the acceleration profiles for the GWL at different depths
are shown together with the acceleration profile for GWL=
-3m depth. The decrease in the water level leads to a
progressive increase in the peak acceleration, varying from
0.38 m/s2 for GWL= 0m depth to 0.42m/s2 for GWL= -3m
depth. This tendency is likely to be related with the reductionof the saturated region, that causes a more significant
transmission of the dynamic motion to the upper elements
located at the level of the embankment. Conversely, after the
peak acceleration is reached (i.e. around 2sec) theacceleration values are smaller for the cases where GWL islocated at higher depth.

loading) and initial effective stress has become very small,which means that liquefaction has occurred.

Horizontal Acceleration (m/s )4 RESULTS AND DISCUSSION

Fig.8 a) to c) shows the 2D representation of the final values(t=20sec) of ESDR for GWL’s located at different depths.
ESDR values closer to 1 it suggest that the ratio between thecurrent effective stress (which is decreasing during cyclic

175

0.4

GWL= 0 m depth
GWL= -1 m depth
GWL= -3 m depth

0.2

0.0

-0.2

-0.4
0

2

4

6

8

10

12

14

16

18

20

Time (sec)Figure 7: Top horizontal acceleration at the top boundary, t=20sec(GWL= -3m, surface).

displacement reduction of 24% compared with a marginalincrease of 4% obtained for GWL located at -3m depth. The
peak acceleration values were higher for GWL depths locatedat deeper levels motivated by the transmission of dynamic
motion to the upper unsaturated elements (that typicallybehaved like dry soil). However, the order of this difference
was found to be small.
This study shows that artificially GWL variation can be used
as an effective method to control the magnitude of thesuperstructures deformations caused by the occurrence ofliquefaction. In this study a 1Hz sinusoidal input wave wasadopted for simplicity. Similar results are likely to beobtained for a real earthquake wave, on which frequency andacceleration contents are quite different. Further studies using
real earthquake data are necessary to confirm the present
findings. Moreover, the successful application of thismethodology lies on an adequate GWL numerical study, sothat the best cost-benefit ratio (GWL depth vs. verticaldisplacement reduction) can be found.

(a ) Case 1: GWL = 0.0m(b ) Case2: GWL = -1.0m

6 ACKNOWLEDGMENT(c) Case3: GWL = -3.0m
Figure 8: 2D representation of ESDR values at t=20sec.

Overall, the increase in GWL depth seems to decrease thearea that is affected by liquefaction (i.e. ESDR =0.99).
In all the cases (Fig. 8a to 8c) ESDR values obtained for theunsaturated domain were smaller or even negative,
particularly on top of the embankment. In this domain there isnearly no increase in pore water pressure and consequently
the effective stress does not decrease during cyclic loading.
However, negative ESDR values (i.e. ESDR=-0.471) indicatethat effective stress is increasing beyond its initial value thisis likely to be related with the definition of effective stress inthe unsaturated domain; where matric suction (negative by
convention) plays an important role in the shear strength of
the soil. Below the embankment levee were obtained smaller
values of ESDR this tendency is related to the effective
confining stresses acting on the ground, which are higher forthe portion below the embankment. Note also that theliquefaction condition closer to the toe of the embankment(Fig 8a) that are likely to cause severe lateral deformations to
the embankment structure are considerably minimised for thecases where GWL in located at a deeper level.

The Authors wish to acknowledge the financial support givenby the Japanese Ministry of Education (Monbugakusho).
Thanks are also due to Dr. Kato, Mr. Morinaka and Mr.
Sanagawa for assisting with the details of some of thenumerical work contained in this study.
7

REFERENCES[1]

Okamura, M. Inexpensive liquefaction countermeasure- soil
desaturation by air injection. Tsuchi to kiso-JGS , 54-7, 2006, pp 28 –
30 (Japanese)
Yoshimi Y., Tanaka, K., Tokimatsu, K. Liquefaction resistance of apartially saturated soil. Soils and Foundations, 29-3, 1989, pp. 157-162.
Nakazawa, H., Ishihara, K., Tsukamoto, Y. and Kamata, T. Casesstudies on the evaluation of liquefaction resistance of imperfectly
saturated soil deposits. In Proc. Of Cyclic Behavior of Soil andLiquefaction Phenomena Conference, Triantafyllidis(ed), 2004, pp.
295-304.
Oka, F.,Yashima, A., Taguchi, Y. & Yamashita, S. –Geotechnique,
49-5,1999, pp.661–680.
Oka, F., Kodaka, T., Kimoto, S., Kato, R. & Sunami, S. KeyEngineering Materials, 340-341, 2007, pp.1223–1230.
Kato, R.; Oka, F., Kimoto, S.; Kodaka, T. & Sunami,S. Transaction of
JSCE, C, vol65-1, 2009, pp.226-240
The LIQCA Research and Development Group – User’s Manual for
LIQCA2D04, 2005

[2]
[3]

[4]
[5]
[6]
[7]

5 CONCLUSION
An alternative cost-effective method for liquefaction
mitigation inducing GWL variation was investigated. The
artificially variation of GWL depth causes the soil to bepartially saturated and in turn strengthens the soil againstliquefaction hazards. In fact, smaller vertical displacements
were observed for the cases where GWL was located at
higher depth. Furthermore, even a small variation of GWL, i.e.
GWL=-1m depth was enough to produce a vertical176

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Development of a New Type Reinforced Soil WallHara, T. Tsuji, S. Yoshida, M. Ito, S. and Sawada, K., Department of civil engineering, Gifu Unicersity



ABSTRACT: Our aim in this study is to develop a new type ofreinforced soil walls as substitute for concrete ones. The structurethat piles are inserted into reinforced soil wall body has been studied
as the new type structure. In this development, a static loading test, adynamic centrifuge test and an impact loading test were carried out
in order to confirm the practicability of the structure to actual diverse
sites. In this paper, the details of the static and impact loading tests
are introduced and the practicability of the structure is discussed.

Keywords: Reinforced soil, pile foundation, retaining wall,
rockfall protection structure
1. INTRODUCTION
The high ductility of soil structures reinforced by geogrid iswell known, as is the possibility of building independent soil
structures. The independent reinforced soil structure, which isreferred to “Geo-wall” in this paper, has been applied to suchdiverse structures as rockfall protection walls [1], mud andsnow avalanche protection walls [2] and the suchlike. Fig. 1
shows an example of the application of Geo-wall to a rockfall
protection wall. Since Geo-wall can be built using existingsoil at the construction site if it is compactable one, they arebeing used ever more frequently as one of economic and CO2
reducible structures.

Therefore, a new type reinforced soil wall that piles areinserted into the Geo-wall body as shown in Fig. 2, which isreferred to “Piled Geo-wall” in this paper, has been developedby authors. The application of piles to Geo-wall in order toimprove lateral resistance is a completely new approach,since piles tend to be vertical bearing piles and are applied tosoil fillings or as reinforcements to soil fillings on soft groundvertical pile [3]. In this novel approach, an assumption ismade regarding the interaction between the pile and Geo-wall,as shown in Fig. 3. The high ductility of the Geo-wall was
assumed to make it possible to transmit lateral force to thepiles despite large relative displacement between the pile andGeo-wall. The validity of this assumption and practicabilityof Piled Geo-wall to actual diverse structure were confirmed
through three experiments, a dynamic centrifuge model test[4]-[6], a static loading test and an impact loading test.
Corroboration of Pile andreinforced soil wall
(Piled Geo-wall)Fig. 2. An image of Piled Geo-wall.

Force
P

Pile foundation

Soil structure(with geogrid)Fig. 1. Application to a rockfall protection wall.
At present, however, the adoption of the spread foundation
for Geo-walls makes the design too wide for application to
narrow construction sites, such as beside mountainous road.
If a narrow Geo-wall as like as RC structure with pilefoundation is achieved, Geo-walls could be widely applied.
And it can also be applied as substitute for concrete structuresand contribute sustainable development.

Soil structure(without geogrid)
Displacement
Fig. 3. Interaction between pile and Geo-wall.
In this paper, the details of the static and impact loading tests
are introduced and the practicability of the Piled Geo-wall is
discussed.

177

2. STATIC AND IMPACT LOADING TEST
2.1 Geo-walls for Experiments
Three full scale Geo-walls, normal Geo-wall (without pile),
Piled Geo-wall-1 (PGW-1) and Piled Geo-wall-2 (PGW-2)
were built on an actual field, and, at first, a static loading test
was carried out with using the normal Geo-wall and PGW-1
and then an impact loading test was carried out with using all
Geo-walls as shown in Fig. 4. Respective structural
conditions, measurements and loading conditions areillustrated in each section of experiment. Fig. 5 shows theappearance of the Geo-walls.
Normal Geo-wall

Piled Geo-wall-1

Piled Geo-wall-2

Impact loading

Impact loading

Impact loading
PD-2

PD-1

Impact loading test
Static loading

2.2 Ground Condition
The foundation ground consists of three types of sandy silt,
soft buried one, medium stiffness and stiffness ones. Fig. 6
shows the results of dynamic penetration test converted to N
value of standard penetration test (SPT). The location of theborings, PD-1 and 2, is shown in Fig. 4. The thickness of thesoft buried top soil is varied from the place of normal
Geo-wall (PD-1) to Piled Geo-walls (PD-2).
2.3 Static Loading Test
A static loading test was carried out in order to confirm thepracticability of Piled Geo-wall to earth retaining wall. In thistest, two full scale Geo-walls, PGW-1 and normal Geo-wall(without pile), were adopted. Fig. 7 shows the structuralsummary of the test. Static horizontal load are given by two
jacks set at 1.0m in height. And the horizontal load is carried
as distributed pressure through an EPS of 0.5m in thickness, asteel plate of 0.12m in thickness and H steel (H200 x 200 x 8x 12).

Static loading

Cross 断面図
section1000 500

Static loading test

Cross断面図
section1000 500

H200X200
Jack (50t)
Load cell

2000

2000

Fig. 4. Location of Geo-walls and respect experiments.

H200X200
Jack (50t)
Load cell

Reaction wall

Normal Geo-wall

Reaction wall

Plate
EPS

Plate
EPS

4000

Piled Geo-wall-1 (PGW-1)
Piled Geo-wall-2(PGW-2)Longitudinal additional geogrid
帯状補強材
H300X300

Plane
view
平面図

EPS
Plate
H200X200
Reaction wallPlane
view EPS
平面図

Plate
H200X200
Reaction wallFig. 5. Appearance of Geo-walls.

Unit: mmN-value of SPT (converted)
１0cm毎の換算Ｎd値
（回）
0

10

20

30

40

(a) PGW-1(b) Normal Geo-wallFig. 7. Structural summary of static loading test.

50

0

深
さ （GL- m）
Depth

1

PD-1
PD-2

2
3
4
5

Load cell
3000

3000

Load cell

The procedure of Geo-walls’ building is as follows;
Pile installing (Piled Geo-wall only): H steel piles(H-300x300x10x15) of 6.0m in length are installed in 4.0m
into the ground.
Setting of steel face member and installing of geogrid ateach layer: steel face members were set on the both side of
Geo-wall and a geogrid was installed at one layer. In theconstruction of Piled Geo-wall, a geogrid with holes locatedin the piles was installed through the piles, as shown in Fig.8.
Fig. 9 shows the tensile stiffness of the geogrid.

6

Fig. 6. Results of dynamic penetration tests.
178summary of measurements.

Steel facememberElevation

Cross section(b)(b)Additional
Additionalgeogrid
geogrid

Plane view

Geogrid： strain gauge (horizontal geogrid)： strain gauge (additional geogrid)lapFug. 8. Installing of steel face member and geogrid.
(a) Horizontal
Horizontal installed(a)installedgeogrid
geogrid

3.0
Tensileforce (kN/m)
引張力

Fig. 11. Measurements of strain in geogrids.

実験で用いる補強材

2.5

Elevation

Cross section

2.0
Elevation

実際の補強材

1.5

①

②

③

④

⑤

⑥

⑦

⑧

⑨

1.0
0.5
0.0
0

10
20
Strain (%)
ひずみ
(%)

30

Fig. 9. Tensile stiffness of geogrid.

(b)Wall
Wall deformation
deformation(b)
Plane view

Installing of longitudinal additional geogrid at eachlayer (Piled Geo-wall only): the additional geogrid was
installed on the loading side of piles, as shown in Fig. 10, in
order to transmit the load to the piles from the Geo-wall bodysmoothly.

： earth pressure meter： strain gauge (pile)(a)PilePilestrainstrain and earth(a)earthpressurepressure
Fig. 12. Measurements of strain of pile, earth pressure anddeformation of Geo-wall.
転倒（主働）モーメント(kN･m)
Falling moment (kNm)

400

Longitudinal
additional geogrid

Falling moment of corresponding to 5.0m heightretaining wall
5ｍ擁壁相当350
300

Piled Geo-wall：杭あり
Normal Geo-wall
：杭なし

250
200

4.0m
height wall
4ｍ擁壁相当150
100

3ｍ擁壁相当3.0m
height wall50

2ｍ擁壁相当2.0m
height wall0
0

Fig. 10. Installing of longitudinal additional geogrid.

30

60

90

120

150

Displacement (mm)

変位 (mm)Fig. 13. Response of Geo-walls.
The executions of each layer, 0.5m in thickness, wererepeated until that the Geo-walls, 2.0m in height, werecompleted.
The measurements of strains of geogrids, additional
longitudinal geogrids and a pile, earth pressure of the
Geo-walls at the loading side, deformation of the Geo-walls
and jack pressures were planned. Figure 11 and 12 show theFig. 13 shows the relationship between falling moment anddisplacement of center top of the Geo-walls (No.2 of Fig.
12(b)). Where, in order to compare with the active earthpressures loaded to retaining walls with different height,
vertical axis is presented by falling moment. According to theresults, high resistance characteristics of the Piled Geo-wall

179

6

6

5

5

350
300

4

Pile length
高さ (m)(m)

3

2

0
20

40

150
Geo-wall
Pile

100

0
3

30

60

90

120

Displacement (mm)Fig. 16. Response of pile and Geo-wall body.
2

2.4 Impact Loading Test
An impact loading test was carried out in order to confirm thepracticability of Piled Geo-wall to rockfall protection walls.
In this test, three actual scale models, the two models of thenormal Geo-wall and PGW-1, which are adopted in the static
loading test, and a new one (PGW-2) were adopted, as shownin Fig. 4. The piles of PGW-2 are installed at outside of
Geo-wall as shown in Fig. 17, are adopted.

0
0

200

0

1

1

250

50

Ground surface

4

Pile length
高さ (m)(m)Fig. 16 shows the responses of the pile and the Geo-wall bodyat the point of pile top (No.1 of Figure 12(b)). According to
the results, large relative displacement of the pile andGeo-wall is occurred though, the lateral force is transmitted tothe pile from the Geo-wall and the lateral resistance of the
Geo-wall was improved as abovementioned. Namely, thevalidity of the assumption established previously, “Geo-wallwith high ductility can transmit lateral force to piles despitelarge relative displacement between the pile and Geo-wall asshown Fig. 3”, could be confirmed.

Horizantal load (kN)can be confirmed from the view point of that the proof
strength of Piled Geo-wall is still increasing even if
corresponding earth pressure of retaining wall of 4.0m inheight is loaded. Against the Piled Geo-wall, because of soft
bearing foundation shown in Figure 6, the normal Geo-wallindicated the tendency of falling by corresponding earthpressure of retaining wall of 2.0m in height. Namely, in caseof adoption of normal Geo-wall in this site, too wide
Geo-wall has to be designed.
Fig. 14 shows the maximum response distribution of pile.
According to the results, it can be confirmed that thehorizontal load is transmitted to the pile and the pile
contributes for improving the lateral resistance of the narrow
Geo-wall.

60

0

50

100

150

曲げモーメント(kNm)
Bending
moment(kNm)

変位 (mm) (mm)Displacement(a) Displacement(b) Bending moment
Fig. 14. Maximum response distribution of pile.

Elevation

Cross section

Fig. 15 shows the relationship between horizontal load andstrain occurring in longitudinal additional geogrid. According
to the results, the strain occurring in the geogrid of the pileside is larger than the center one and the strain depends on theintensity of the horizontal load, thus the geogrid is consideredto sufficiently function for transmitting of lateral force topiles from Geo-wall body.
400

vicinity of pile
center of span

Horizontal load (kN)

350

H300X300

300
250
200
150
100

Plane view50

： strain gauge (geogrid)： strain gauge (pile)

0
0

0.3

0.6

0.9

1.2

1.5

Strain (%)

Fig. 15. Strain occurring in additional geogrid.

Fig. 17. Structural condition and measurements of PGW-2.
180

PGW-2 was adopted to confirm the best position of pile
installing in case of impact loading, because there considereda possibility that keeping wide energy absorption extent isbetter than receive the rockfall energy at the center of
Geo-wall body as PGW-1. In addition, the loading directionof static and loading tests are different, thus longitudinal
additional geogrid of PGW-1 was installed on both sides of
the Pile previously. In PGW-2, the additional geogrid was notinstalled. In this impact test, the impact of 100kJ was loaded
to respective Geo-walls as shown in Fig. 18.

Fig. 19. Normal Geo-wall.

Iron ball (9kN)

Geo-wall
100kJ impact(a) Loading side

11.1m

Fig. 18. Impact loading test.
Fig. 19 to 21 show the respective test results.
According to the results, against the collapse of normal
Geo-wall, both Piled Geo-walls could receive the impact
energy. Namely, the effectiveness of application of piles toimprove lateral resistance of Geo-wall could be confirmed.
Note, however, that the deformation of the back side of the
Geo-wall (protecting side, road side for instance), one of
PGW-2 (piled installed at outside of the Geo-wall) is smaller
than one of PGW-1 (piles installed at center of the Geo-wall).
Since this result, the possibility that the best position of pile
installing is outside of the Geo-wall in impact loading casecan be considered. Meanwhile, from Fig. 22 showing
responses of piles of the PGW-1 and 2, it is also consideredthat the resistance of the piles of PGW-1 seems to be smaller
than ones of PGW-2. With respect to the result, following
reasons can be considered.
- There might be a void in front of resistance surface of thepiles of PGW-1 before the impact loading test because of thestatic loading test implementation.
- In the foundation ground that the piles of PGW-2 were
installed, the soft buried soil is comparatively thin, because
the piles of PGW-2 were installed at the position closed tovicinity hill of 0.65m.
The best position of pile installing in case of impact loading
has to be studied again with considering abovementioned
phenomenon in the near future.

(b) Back side (protecting side)
Fig. 20. Piled Geo-wall-1 (PGW-1)(a) Loading side(b) Back side (protecting side)
Fig. 21. Piled Geo-wall-2 (PGW-2)
181

Height from pile tip (m)6Ground surface

4

PGW-2 (Max)PGW-2 (Res)2
PGW-1 (Max)PGW-1 (Res)
0
0

5

10

15

20

25

Displasment (mm)(a) Displacement

Height from pile tip (m)6PGW-2 (Max)PGW-2 (Res)PGW-1 (Max)PGW-1 (Res)Ground surface

4

2

0
0

20

40

60

80

100

120

Bending Moment (kNm)(b) Bending moment
Fig. 22. Response of piles of PGW-1 and 2.
Fig. 23 shows transition of strain occurring in the longitudinal
additional geogrid and the pile of PGW-1. According to theresults, from the view points of the strain occurring in thegeogrid and the timing of occurrence of the strain between thegeogrid and the pile, effect of the geogrid to transmit thelateral forces to the piles from Geo-wall could also be
confirmed in the impact loading test.
Strain (με)

20000vicinity of pile

center10000

carried out in this study.
Unification of piles and Geo-wall body: Geo-wall with highductility can transmit lateral forces to piles despite large
relative displacement between the pile and Geo-wall. And
adoption of longitudinal additional geogrid is effectiveness toachieve the unification of piles and Geo-wall body. Note,however, that the effect of the relative displacement to theinternal quality of the Geo-wall body has to be studied.
Application to retaining walls: High resistancecharacteristics against static load and dynamic earth pressure
could be confirmed from the static loading test as well as theprevious carried out dynamic centrifuge model test [4]-[6].
Therefore, application of Piled Geo-wall to retaining walls is
sufficiently possible for all ones depending on the tolerable
deformation of the Piled Geo-wall and the embankment
sustained by Piled Geo-wall.
Application to rockfall protection walls: High performanceof Piled Geo-wall to receive rockfall could be confirmed fromthe impact loading test. Hence, application of Piled Geo-wallto rockfall protection walls would be possible if it is similartarget rockfall with the experiment carried out in this study.
Further experiments to determine the best position of the pile
installing and to study the design for appropriate size of Piled
Geo-wall depending on the intensity of rockfall energy arenecessary before the widely application.
Design code for the application: Studies on the design
method based on numerical simulations of the test results are
conducting by authors. Although needed further experiments
and studies to solve several issues are remained asabovementioned, a valuable design code for widely use of
Piled Geo-wall would be published in the near future.
Expectation of Piled Geo-wall to sustainable
development: Piled Geo-wall has possibility to be used asalternative structure of concrete ones with similar size.
Therefore, if the development of Piled Geo-wall is achieved,
it can be expected that the Piled Geo-wall contribute tosustainable development from the viewpoint of reducible
structure of CO2.

0-10000

0

1

2

3

4

5

4. REFFERENCES

Time (sec)[1]Strain (με)(a) Additional geogrid
-50 0
-150
-250

1

2

3

GL-3.0m
GL-0.7m
GL+1.4m

GL-2.1m
GL+0.0m

4

5[2]

GL-1.4m
GL+0.7m[3]-350Time (sec)(b) Pile
Fig. 23. Strain of additional geogrid and pile of PGW-1.

[4]

[5]

3. CONCLUSION AND DISCUSSION ON
PRACTICABILITY OF PILED GEO-WALL[6]

Application of piles to Geo-wall: The effectiveness ofapplication of piles to Geo-wall in order to improve lateralresistance of Geo-wall was confirmed from all the tests
182

Yoshida, M., Tatsuta, N., Nishida, Y., Inoue, S., Yashima, A., Sawada,
K. & Moriguchi, S., “Full-scale field test of reinforced embankment
adjacent to steep-slope,” Geosynthetics Engineering Journal, Vol.20,
2005, pp. 295-300.
Itou, S., Yokota, Y., Kubo, T. & Arai, K., “Field loading test of theprotection embankment retaining wall reinforced with geosynthetics,”
Geosynthetics Engineering Journal, Vol.15, 2000, pp. 340-349.
Hara, T. & Tsuji, S., “A reinforced soil structure with pile foundation,”
Geosynthetics Engineering Journal, Vol.23, 2008, pp. 209-214.
Hara, T., Tsuji, S., Yashima, A., Sawada, K., Tatta, N., “Dynamic
interaction between pile and reinforcement soil structure – Piled
Geo-wall –,” in Proc. of Int. Conf. on IS-KYOTO, 2009, pp. 457-463.
Hara, T., Tsuji, S., Yashima, A., Tatta, N., “Dynamic centrifuge model
test on Piled Geo-wall,” Geosynthetics Engineering Journal, Vol.24,
2009, pp. 257-262.
Hara, T., Tsuji, S., Yashima, A., Sawada, K., Tatta, N. & Otake, Y.,“Independent reinforced soil structure with pile foundation,” J. of
SOILS AND FOUNDATIOS, Vol.50, No.5, 2010, pp. 565-571.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Development of Design Charts for Reinforced Embankment Using Excel
Spreadsheet Slope Stability Analysis
K. Ito, M. Z. Hossain and T. Sakai, Graduate School of Bioresources, Mie University



ABSTRACT: The present study is an attempt to develop design
chart for reinforced embankment using excel spreadsheet slope
stability analysis. The selected reinforcing materials that have been
identified for reinforced embankment are composite materials withsmooth surface and rough surface made of used cells, stones, wood,concrete and brick. Utilizing simplified available method andequations, the design charts for reinforced embankments have beenpresented using simple techniques. The most critical slip surface has
been identified based on random searching that produced minimum
factor of safety. Results for different slope angle with varying
reinforcing layers and spacing corresponding to the most critical slip
surface are given. It is expected that this study will facilitate indesigning and constructing reinforced embankments containing
various types of reinforcing materials.

2. METHODOLOGY
There are several methods for stability analysis of slopes[2-15]. Any method can be used to solve the equilibrium
equations of slices and to search for the critical slip surface
automatically, by the combined use of spreadsheet iteration
method. Here, Jambu’s generalized procedure of slices isused [16]. Equations for generalized procedure of slices are
given as follows (Fig.1). Fill embankment considered in thepresent study is shown in Fig.2.

Keywords: Reinforced Embankment, Slope Stability,
Composite Reinforcements, Design Charts
1. INTRODUCTION
Compared to conventional soil reinforcing material,
composite materials generally exhibit superior behavior in
performance and strength. Generally, the conventional
reinforcements, used for reinforcing soil, contain only one
type of material such as geogrid, geosynthetic or wire meshetc. It is known that the material used in soil reinforcement
applications must be safe against tension failure and adhesion
failure for its effective utilization in the field and reliable
design of earth structures. For a given situation, single type ofmaterial can provide limited reinforcement capability inreinforced soil structures due to its low frictional resistanceand poor cohesion. For an optimal response, therefore,
different types of reinforcement, that fulfills both therequirements such as possess adequate tensile strength andadequate frictional resistance, is getting considerable
attention lately. Thin-reinforced-mortar composite consistingof evenly distributed fine mesh as the reinforcement andcement-sand mortar as the matrix showed enhanced
performance because of its synergetic action of mesh withmortar and mortar with soil [1]. In view of the above distinct
advantages of composite material, a study is undertaken withthe following objectives; a) to present a simplified technique
for slope stability analysis using excel spreadsheet and b) todevelop design charts for reinforced embankment containing
different composite materials in order to facilitate safe andreliable construction using locally available materials.

p  z  q PxFig.1 Symbol used in Janbu’s generalized procedure of slices
16
14
12
10
8
6
4
2
0
0

5

10

15

20

25

30

35

Fig.2 Example of search of critical slip surface
For Vertical equilibrium
(1)
For
(2)
For
(3)

183

(12)

For
.

(13)

(4)
(5)
.

(14)

(6)
(15)
(7)
(16)
(8)
(17)

(9)

(18)

(10)

3. SPREADSHEETMETHOD

AUTOMATEDITERATION(11)

The spreadsheet approach is shown in Table 1 as an example
of 19 slices.

Table 1. Input fundamental properties of slopes
X
Slice#

Y

Ytp

hij

Ea = 0

9.5

15

15

0 tan a

DX

P

Bo

1

10.5

11.4

15

3.6

3.6

1

2

11.5

9.5

15

5.5

1.9

1

81.9

155.6149.14 0.362396 135.5974 77.38782 125.0832

3

12.5

8.3

15

6.7

1.2

1

109.8

131.7665.88 0.583429 112.9186 66.62053 191.7037

4

13.5

7.4

15

7.6

0.9

1

128.7

115.8377.22 0.727997 106.0718 54.64025

5

14.5

6.7

15

8.3

0.7

1

143.1

100.1785.86 0.836967 102.5847 40.99186 287.3358

6

15.5

6.1

15

8.9

0.6

1

154.8

92.88

92.88 0.891017 104.2404 32.74674 320.0826

7

16.5

5.6

15

9.4

0.5

1

164.7

82.35

98.82 0.941189 104.9949 21.78152 341.8641

8

17.5

5.25

15

9.75

0.35

1

172.3560.3225103.41 1.000927 103.3143 0.723512 342.5876

32.4

A'o
116.64

n a0

Ao

D Eo

0

19.44 0.162658 119.5148 47.69536 47.69536

9

18.5

4.95

14

9.05

0.3

1

169.2

50.76

1.01458 100.0612-6.96237 335.6252

10

19.5

4.8

13

8.2

0.15

1

155.2523.2875

93.15 1.029776 90.45657-28.8943 306.7309

11

20.5

4.65

12

7.35

0.15

1

139.9520.992583.97 1.029776 81.54201-26.0467 280.6842

12
13
14
15
16
17
18
19

21.5
22.5
23.5
24.5
25.5
26.5
27.5
28

4.6
4.7
4.9
5.25
5.7
6.25
6.95
7.5

11
10
9
8
7.5
7.5
7.5
7.5

6.4
5.3
4.1
2.75
1.8
1.25
0.55
0

0.05
-0.1
-0.2
-0.35
-0.45
-0.55
-0.7
-1.1

1
1
1
1
1
1
1
0.5

123.75
105.3
84.6
61.65
40.95
27.45
16.2
4.95

6.1875-10.53-16.92-21.5775-18.4275-15.0975-11.34-2.7225

74.25
63.18
50.76
36.99
24.57
16.47
9.72
1.485

-36.0076-48.688-49.6864-48.9061-38.6898-30.4538-22.4364-5.81733

Ytp=I=Xi/slope ratio P=J=1/2(hij1+hij2)γ184

101.52

246.3441.015111
0.955151
0.893659
0.780810.699511
0.618707
0.505315
0.276801

73.14472
66.14659
56.80018
47.37386
35.12452
26.62005
19.23553
5.3648630-48.6881-47.90612-28.4538
3-2.81733

Table 2 Calculation of factor of safety
S B1
646.5529

Calculation of T1

S A1
1563.479

Col( 15 ) Col( 16 )DEo / Dx tana tCol( 17 )ht

Col( 18 )T1

Col( 19 )D T1

Col( 20 )t1

Col( 21 )B1

Col( 22 )A'1

Col( 23 )na1

Col( 24 )A1

62.54159
72.00417
60.63039
47.81606
36.869327.26413
11.25251-3.11943-17.9283-27.4705-31.0271-42.3478-49.1872-49.2962-43.798-34.5718-26.4451-18.8358-11.6347

0
1.2
1.833333
2.233333
2.533333
2.766667
2.966667
3.133333
3.25
3.016667
2.733333
2.45
2.133333
1.766667
1.366667
0.916667
0.6
0.416667
0.183333
00-39.419-26.431-17.9551-26.6724-32.085-47.1494-78.6968-90.0753-233.084-208.003-197.646-90.342-73.9139-67.5715-35.3575-20.4764-21.4519-2.05324-2.06604-39.41912.98792
8.475943-8.71727-5.41262-15.0645-31.5474-11.3784-143.009
25.08112
10.35643
107.3043
16.42806
6.342401
32.21401
14.88109-0.97545
19.39864-0.01281-39.41912.98792
8.475943-8.71727-5.41262-15.0645-31.5474-11.3784-143.009
25.08112
10.35643
107.3043
16.42806
6.342401
32.21401
14.88109-0.97545
19.39864-0.02562-25.2682
180.287
141.9311
107.9845
96.38117
83.84132
66.57632
56.34005
7.857412
27.04967
22.54596
11.55272-12.1728-18.1885-32.8524-25.124-14.561-24.919-2.70841-4.21137
56.93275
70.96557
71.98964
82.61243
83.84132
79.89158
96.58294
15.71482
108.1987
90.18386
138.6326
73.03684
54.56544
56.3184
33.49865
15.88473
21.35918
1.477315

0.084389
0.237306
0.434163
0.577081
0.694379
0.757116
0.819786
0.906292
0.931045
0.985251
0.985251
0.999973
0.985202
0.952026
0.875446
0.81309
0.746867
0.647903
0.427869-49.9043
239.9127
163.4539
124.7478
118.9731
110.7377
97.45425
106.5694
16.87869
109.8183
91.53385
138.6363
74.133957.31506
64.33113
41.19917
21.26847
32.96664
3.45273

2.4
1.266667
0.8
0.6
0.466667
0.4
0.333333
0.233333
0.533333
0.433333
0.433333
0.366667
0.266667
0.2
0.1-0.13333-0.36667-0.46667-0.73333

Fs =2.418176
Col( 25 )D E1

Col( 26 ) ts
E1(Ea = 0)0-4.63107 -4.63107 -1.47831 -1.69705
81.0748 76.44374 21.52109 53.99784
74.33725 150.781 27.70241 85.03305
56.39691 207.1779 28.50141 94.33146
47.18165 254.3595 33.01981 114.5735
38.04744 292.407 33.67198 119.5324
26.2756 318.6826 32.24058 117.0323
12.2699 330.9525 39.26071 147.2303
0.877485331.83 6.403602 24.27029-18.364 313.4659 44.41438 173.669-15.3065 298.1595 37.01949 144.7535-45.7782
0 57.18797 228.1949-42.8298 -42.8298 30.35341 124.7634-41.8903-84.72 22.79017 95.50043-59.4556
1 23.69992 102.159-42.1613 -41.1613 14.16822 62.20679-23.3563 -64.5175 6.752594 30.18848-38.5519
2 9.149566 42.00334-4.13623 -2.13623 1.292148 6.345746

In the field of earth slope protection, the basic materials
needed for cement composite constructions are wire mesh,sand, cement and water. The skeletal steel is used in theconstruction of cement composite frame structures for earth
slope protection. A basic description of the constituent
materials and construction procedure is given below. After
iteration, the solver automatically searches the critical slip
surface as shown in light blue color or orange color in Fig.1
and determines the minimum factor of safety [FS less than 1
depending on slope]. Reinforcements for this critical slip
surface having minimum FS needs to be calculated for thesafe design and construction of embankments.
4. RESULTS AND DISCUSSION
Reinforcements for embankments having slope angle 45 and
60 are calculated and shown in Fig.3. The FS values are
greater than one by inserting 3 layers of reinforcements[Fig.4].

Fig.3b Reinforced slope with 45 angle

Fig.3a Reinforced slope with 60 angle
Fig.4 FS vs. types of reinforcements with 3 layers
185

It is observed that the embankments with 60 slopes are
giving FS values of more than 1.5 for reinforcements madeof wood, concrete and brick only. The embankments with 45
slopes giving lower FS for all the cases. This type of
behavior was absolutely obscured in the case of Fig.5
although the numbers of layers were same. This factor needs
to be carefully considered during construction reinforced
embankments.

For safe design, the FS should be more than 1.5 whichindicates that the number of layers should be more than threelayers. For economic design, it is necessary to minimum
number of layers of reinforcements. Considering the bothsafe and economic design, the number of layers wereincreased to 4 and FS has been calculated and given in Fig.5.
It was observed that the FS values are greater than 1.5 forreinforments containing stones, wood, concrete and brick.

5. CONCLUSIONS
Design charts for reinforced embankments containing
various types of reinforcements have been presented in thispaper. In the process, the powerful spreadsheet techniques
are used to render the iterations in the search for critical slip
surface and for calculation of minimum factor of safety. A
noteworthy feature of the proposed study is that the search
for the critical slip surface is non-circular and automatic,
despite the fact that the FS is non-explicit. It is expected thatthe present study will bring the new concept for the design ofreinforced slope. The conceptual simplicity, ease of
implementation and practical versatility of the proposed
perspectives and spreadsheet methods are likely to apply
also to other reinforcement and slopes.
6. REFERENCES
Fig.5 FS vs. types of reinforcements with 4 layers at bottom
part[1] M. Zakaria Hossain, “Pullout Response of Cement composite Members
Embedded in Soil,” ACI Materials Journal, V. 105, No. 2, March-April
2008, pp.116-124[2] Janbu,N., “Slope stability computation. In Embankment-Dam
Engineering,” Casagrande Volume,ed.R.C. Hirschfeld and S. J.
Poulos .Krieger Pub. Co.,1987,pp.47-86[3] Bishop, A.W., “The use of the slip circle in the stability analysis ofslopes.” Geotechnique, 1955,5,7-17[4] Morgenstern, N.R. and Price,V.E., “The analysis of the stability ofgeneral slip surface. Geotechnique,1965,15,79-93[5] Spencer, E., “A method of analysis of the stability of embankments
assuming parallel inter-slice forces.” Geotechnique,1967,17,11-26.
[6] Chen,Z. and Morgenstern, N.R., “Extensions to the generalized methodof slice for stability analysis.,” Canadian Geotechnics Journal, 1983, 20,
104-119[7] Sarma,S.K., “Note on the stability of slopes.” Geotechnique,1987,37,
107-111.
[8] Hoek,E., “General two-dimensional slope stability analysis.In
Analytical and Computational Methods in Engineering RockMechanics,” Ed. E. T. Brown.Allen & Unwin,1987.pp.95-128.
[9] Wright,S.G., Kulhawy,F.G. and Duncan, “J.M.,Accuracy of equilibriumslope stability analysis.Journal of Soil Mechanics and Foundations
Division,” ASCE,1973,99,783-791.
[10] Fredlund, D.G.and Krahn,J., “Comparsion of slope stability methods ofanalysis,” Canadian Geotechnical Journal,1977,14,429-439.
[11] Skempton, A,W. and Vaughan, P.R., “The failure of Carsington dam,”Geotechnique,1933,43,151-173.
[12] Alonso, E.E., Gens, A. and Lloret, A., “The landslides of Cortes dePallas, Spain,” Geotechnique,1933,43,507-521.
[13] Wright,S.G. and Duncan, J.M., “Limit equilibrium stability analyses
for reinforced slopes,” TRB Paper No.910441,Transportation Resarch
Board, Washington DC,1991.
[14] Chen,Z. and Shao,C., “Evaluation of minimum factor of safety in slope
stability analysis,” Canadian Geotechnical Journal,1988,25,735-748.
[15] Ugai,K. and Leshchinsky,D., “Three-dimensional limit equilibrium
and finite element analyses:a comparsion of results,” Soils andFoundations,1 995,35,1-7.
[16] Duncan, J.M., “State of the art: limit equilibrium and finite-element
analysis of slopes, “Journal of Geotechnical Engineering, ASCE,1996,
122, 577-596.

The above FS are obtained when the reinforcements areinserted at the bottom part of the embankment. This was
performed for the ease of calculation. However, for the easeof construction, it is necessary to insert the reinforcement
uniformly throughout the embankment. Considering thefacts above, the FS with 4 layers inserted at 1m interval arecalculated and given in Fig.6.

Fig. 6 FS vs. types of reinforcements with 4 layers at uniform
distribution
186

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Numerical Simulation of 2D Crack Growth with Frictional Contact in Brittle
Materials
Kai Zhang.1, 2 Jing-cai Jiang.2 Qing Yang.1(1. Institute of Geotechnical Engineering, School of Civil Engineering, Dalian University of Technology, Dalian,
China)(2. Department of Civil and Environmental Engineering, The University of Tokushima, Tokushima, Japan)

ABSTRACT: In this paper, the extended finite element method(XFEM) is applied in modeling the 2D crack growth with frictional
contact under uniaxial compress load in the rock-like materials. First,
the implementation of XFEM is incorporated into a commercial
FEM software (ABAQUS) in which the constitutive law of linear
elasticity and the criterion of maximum tangential stress (MTS) is
adopted. Then a user subroutine is coded and incorporated into
ABAQUS to simulate the growth of wing crack with the frictional
contact in the crack faces. A series of numerical simulations of 2D
plane strain rectangle with central pre-set crack are carried out, andcomputed results are compared with experimental ones. The effects
of inclination and coefficient of the friction of the pre-set cracks on
growth of wing cracks are examined. In addition, size effect ofmaterials is also investigated, and these jobs contribute to theunderstanding of 2D crack growth.

Keywords: Extended finite element method (XFEM);
rock-like materials; crack growth; frictional contact; size
effect.
1. INTRODUCTION
The discontinuities (e.g. joint, crack, and void) which exist
previously or appear during the evolution often control the
instability and final failure in the brittle materials. In recent
years, a lot of experiments and numerical simulations havebeen carried on crack growth in brittle materials, such as rock,concrete and so on. It has been widely acknowledged that
FEM is one of the effective methods used to model the crack
problem because of its good adaptability and scalability. But
conventional FEM has many difficulties in dealing with thestrong discontinuities, which mostly lead to high density
meshes on the crack tip or repeated remeshing as crack grows[1].
The Extend Finite Element Method (XFEM) originallyproposed by Belytschko and Black [2, 3] in 1999, is very
powerful for discontinuous problems in fracture mechanics.
They added discontinuous enrichment function to the finite
element approximation to account for the presence of thecrack. Later, other researchers improved the method, andapplied it in many subjects in fracture mechanics.
Actually, the extended finite element method (XFEM) is anextension of the conventional finite element method based on
the concept of partition of unity. It allows the presence of

discontinuities in an element by enriching degrees of freedom
with special displacement functions. It does not require themesh to match the geometry of the discontinuities. It allows
contact interaction of cracked element surfaces based on asmall-sliding formulation and allows both material andgeometrical nonlinearity [4].
During the numerical analysis, the contact is a typicalnonlinear problem, which not only due to the complicatedmechanical model for the contact surface, but also more
originate from the special discontinuous constraint of contact
surface. The convergent results are difficult to obtain only
when the appropriate friction model is chosen. In thesimulation of the paper, the penalty method is used to deal
with the frictional contact of the crack surfaces during theload process, which can better model the impact on thecontact surface when the wing crack grows.
This paper intends to contribute to the understanding of 2D
crack growth under uniaxial compress load in brittle material
with frictional contact, and a series of numerical modeling are
carried out whose results are compared with experimental
ones [5]. The effects of inclination and coefficient of thefriction of the pre-set cracks on growth of wing cracks are
examined.
2. THE XFEM APPROXIMATIONIn the classical FEM, the approximation to displacement fieldu  x  is expressed as [1]nu h x    N iu i

(1)

i 1

where N i is the interpolated shape function associate with
the node i; u i is the classical vectorial DOF(degrees of
freedom) at node i. For each computational point (x, y) in thefield, N i should satisfy PU (partition of unit):ne N  x, y   1 .
i 1

i

(2)

The XFEM enriched a standard approximation locally on thecrack with discontinuous functions.

187

3. FRICTIONAL BEHAVIORS
3.1 Conventional friction theories
Usually, when the inner surfaces of the crack are in contact,
the shear force will be transmitted as well as the normal force
across the interface. We now consider unilateral contact withfriction on the interface.
The crack faces or the interface are assumed asd  d  d with the unit normal vector to d denoteby n in Fig. 2, and in the later modeling flat ellipse is chosen
which is close to the real situation of rock-like material. We
also introduce the displacement and traction on each face ofFig. 1 An arbitrary crack placed on a meshuh  x    N i  x  ui   H  x  N j  x  a j
iI







jJ



  N k  x    bkl1l1  r ,   
kK1
 l 1

4

(3)

 4

  N k  x    bkl 2l2  r ,  kK 2 l 1
where I is the set of all nodes in the mesh; J is the subset ofnodes which support is intersected by the crack but do notcover any crack tips (e.g. the circled nodes in Fig. 1.);K1 , K 2 is the subset of nodes which support conclude thefirst and second crack tips (e.g. the squared and diamond
nodes in Fig. 1.).
The function H  x  is a “generalized Heaviside” function, in
which the discontinuity is aligned with the crack surface d .

a j is the corresponding additional DOF for the discontinuity.

Fig. 2 The crack faces

The frictional model is easily described by using theappropriate form of the displacement and traction on thecrack faces. The normal components are expressed in thecondition of no contact and in contact:w  w  n  0,
Given a point in the domain, we denote the vectorial distancex between it and the closet point on d . Also, the normal

vector to d is constructed. The function H  x  is thengiven by the sign of the scalar product x  en :1, x  en  0,
H x   signx  en   
 1, x  en  0.

(4)

The set of near-tip functions l r ,   are a set of additional
shape functions which span the exact asymptotic crack-tip
fields for a linear elastic material:  r ,   { r sin  2  , r cos  2  ,
l

 

 

(5)

 
 
r sin   sin   , r cos   sin   }22where r ,   are the local polar coordinates at the crack tip.
l

Except this, bk is corresponding additional DOF for thecrack tip.



the crack: w , t on d and w , t on d [6].





(6A)







t  n  0, t  n  0, t  n  t  n,t   n w   w   n  0

(6B)
(6C)

The tangential component will satisfy the equilibrium:(7)n  t n  n  t n  0 ,so that the standard value of the tangential component can be
represented by













p  n  t n .

(8)

The additional equations depend on whether or not there isfriction. When the friction is idealized using a Coulomb law,
two contacting surfaces can carry shear stress up to the
maximum frictional force before they start sliding relative toeach other:(9)g   t n ,where  is the coefficient of friction.
There is two states (sticking and sliding) and two equations to
be satisfied are:if p  g (stick),(10A)
n  w n  n  w  n n  w  n   n  w  n   0 if p  g (slide).


188



(10B)

Table.2 The parameter of material

3.2 Stiffness method for friction
The stiffness method used for friction is a penalty method that
permits some relative motion for the surfaces called elastic
slip when they should be sticking. While the surfaces are
sticking, the magnitude of sliding is limited to the elastic slip
and adjusted to enforce this condition.
The stiffness method requires the appropriate selection of anallowable elastic slip  i . Using a larger  i makes theCompressive
strength/MPa
38.5

no crack
propagation

may occur. So the value of allowable slip used in thesimulation which works very well must provide aconservative balance between efficiency and accuracy.
The allowable elastic slip is given as [4](11) i  Ff li

Possion’s
ratio
0.14

Interative computation for
current crack configuration

Evaluation of thecrack propagation
criterion (MTS)

where F f is the slip tolerance, and li is the characteristic
contact surface length.

crack propagation
Determination of the crack
growth direction (MTS)

4. NUMERICAL EXAMPLES
4.1 The pre-set of parameters
The models used in the numerical study are rectangle blocks
with dimension 240 mm high and 120 mm wide. A
pre-existing flaw (25 mm long and 0.5 mm wide) withdifferent inclinations and coefficients of friction is set up in
the center of each model. The flaw will close during theloading procedure. The other parameters of the flaw and theparameters of the material see Table. 1 and Table. 2.

Fig. 3 The model of a flaw under compression(2a: the length of the flaw, θ: the initial angle)

1-1
1-2
1-3

Yonng’smodulus/GPa
5.93

Loading step

convergence of the solution more rapid, but brings the bad
accuracy. If  i is chosen very small, convergence problems

No.

Tensilestrength/MPa
3.2

Table.1 The dimension parameters of the flaw
Inclination Coefficient
Inclination CoefficientNo.
α/°of friction μα/°of friction μ
30
0.5
2-2
45
0
45
0.5
3-2
45
0.1
60
0.5
4-2
45
0.3
5-2
45
0.7

Crack extension by a crack
segment with a length

Fig. 4 Algorithm used for crack propagation analyses[8]enables the convergent results, but also makes a good crack
propagation path. During the meshing, the encryption(0.1mm*0.1mm) is made around the crack tip, while normalmesh (5mm*5mm) is chosen in other zones. After eachconverged load increment the crack propagation criterion ischecked. If no crack growth happens, the current crack
configuration remains unchanged and the next loading step isapplied; otherwise a new crack segment will be inserted withsome length (thought as the size of element). Subsequently,
the process will continue with the next loading step [8].
4.3 Numerical results
The main findings from the modeling are as follows:(1) Wing cracks are observed in all the models, which showsmode-I cracking is the main failure mode. Wing cracks
appear first with a very short length at the tips of the initial
crack, and then become wider and longer with the loadincreasing, finally towards the direction of the compressive

4.2 Modeling and Computing
Before the analysis of the program, we set appropriateloading steps and the number of iteration, which not only
189(1) 50 steps(2) 100 steps(3) 200 steps

Fig. 5 The typical propagation path of the wing crack

phenomenon is observed in this modeling.
(3) The size has little effect in the vertical load increasing
when the wing crack grows, so the model with fixed widthand height is able to simulation the real results for different
rock mass’s situation.
(4) As the increase of the coefficient of the friction, the wing
crack propagation will be slowly as the vertical load increase.
Fig.7 also shows the crack faces will close when the length ofwing crack reaches near 5mm. The curves extend mostly notas the line, which shows the friction over the crack face makethe stress around the crack changed, thereby inhabit the
propagation of the wing crack.
(5) The length of the wing crack increases nonlinearly as thevertical load increases, in details that increasing rate of crack
length for vertical load increase at first and then decrease fast
when the length of the wing crack reaches the critical value
which means that the stress around the crack is changed dueto the propagation of the crack. The initial load decreases as
the increase of the inclination of the pre-set crack. The resultsin Fig. 8 show the difference of the inclination produce thedifferent distribution of the normal and tangential stress.

Fig. 6 Curve of the length of wing crack vs.
vertical load/the plane width for μ=0.3

5. CONCLUSIONS
The mechanism of crack propagation was studied using
brittle material with the frictional contact under uniaxial
expression and the relationship between the length of wing
crack and vertical load was recorded and analyzed. A XFEM
technique was used to simulate the propagation of wing
cracks in the specimens, and the obtained results werecompared with the laboratory test one [5], which shows they
were very close to the experimental ones. It proves theaccuracy and practicability of the above method in modelingthe frictional contact behavior.

Fig. 7 Curve of the length of wing crack vs.
vertical load for β=45°6. ACKNOWLEDGEMENTS
This research is supported by the NSFC (Grant No.50979013).
The authors are grateful for these supports.
7. REFERENCES
[1]
[2]
[3]

Fig. 8 Curve of the length of wing crack vs.
vertical load for μ=0.5

stress. The path of wing crack seems to be similar to thehyperbola lines in mathematics, which are discussed in theprevious papers of Yang who is one of this paper’s authors.
(2) The crack in the infinite plate will completely keep open
or close under the loading, while in the finite plate middle
region maybe exist, which means the crack will close partly.
Because of the complex in mathematics, it is difficult toobtain the analytical solution of the problem, and the

[4]
[5]
[6]
[7]
[8]

190

Lu-xian, Li. & Tie-jun, Wang, “The extended finite element method
and its applications,” Advances in Mechanics, vol.35 (1), 2005, pp.
5-20.
Belytschko, T. & Black, T, “Elastic crack growth in finite elementswith minimal remeshing,” Int. J. Numer. Meth. Eng. vol.45 (5), 1999,
pp. 601-620.
Moes, N., Dolbow, J. & Belytschko, T, “A finite element method forcrack growth without remeshing,” Int. J. Numer. Meth. Eng, vol.46 (1),1999, pp. 131-150.
SIMULIA, ABAQUS 6.10: Analysis Users’ Manual. 2010.
Shao-hua, G, “The experimental and theoretical investigation on
fracture of rock-type materials under compressive loading,” Ph.D.
Thesis, 2003.
Dolbow, J., Moes, N. & Belytschko, T, “An extended finite element formodeling crack growth with frictional contact,” Computer Methods in
Applied Mechanics and Engineering, vol.190, 2001, pp. 6825-6846.
Yu-wen, D., Qing-wen, R. & Qin, S. “Extended finite element methodof frictional contact problem,” Journal of Yangtze River Scientific
Research Institute, vol.26 (5), 2009, pp. 45-49.
Dumstorff. P. & Meschke, G. “Crack propagation in the framework of
X-FEM-based structural analyses,” International Journal for Numericaland Analytical Methods in Geomechanics. vol.31, 2007, pp. 239-259.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Interface Behaviour of Basalt Geosynthetic with Sand Using Direct Shear Device
Md. Bellal Hossain, Md. Zakaria Hossain and Toshinori Sakai
Department of Environmental Science & Technology, Mie University, JapanABSTRACT: The suitability of the geosynthetic materials depends
on not only the mechanical properties but also the soil-geosynthetic
interaction properties. Therefore, this paper attempted to evaluate
interface behaviour of basalt type geosynthetic with sand through aseries of direct shear tests. In the tests the applied normal stresses
were 40-160 kPa. The test results reveal that the shear strength of all
sand-geosynthetic interfaces increased with the increasing normal
stresses. The geogrid having small apertures and thin ribs shows the
maximum interface shear resistance whereas the sand-geotextile
interface shows significantly lower shear resistance comparing tothat of sand-sand interface. Strain hardening-softening behaviour isclearly observed for the geogrid interfaces, but the geotextile
interface shows completely different behaviour. It is also observedthat the presence of sand in the apertures of geogrids dominates thedilatancy behaviour of the sand-geogrid interfaces. The averageinterface efficiency of the sand-geogrid interfaces tested in this study
ranges from 0.61 to 0.87.

Keywords: Direct shear tests, Basalt geosynthetics, Interface
efficiency, Toyoura Sand.
1. INTRODUCTION
Geosynthetics have been increasingly used in geotechnicaland environmental engineering for the last four decades.
There are a significant number of geosynthetic types andapplications in geotechnical and environmental engineering.
Common types of geosynthetics used for soil reinforcement
include geotextiles (particularly woven geotextiles), geogrids
and geocells. For practical use of geosynthetics as soil
reinforcement materials, the suitability of the material should
be checked by evaluating not only the mechanical properties
but also the soil-geosynthetic interaction properties.
The pullout and the direct shear tests are the two laboratory
tests commonly used to measure the corresponding mobilizedinteraction resistances. With the advancement of soil
reinforcement techniques, a number of researches on
soil-geosynthetic interaction testing have been performed[1]-[5]. The interface shear resistance of geotextiles against
soil results solely from the shear resistance between thegeosynthetic surface against soil particles because soil
particles are not interlocked with openings. On the other hand,
Geogrids are characterized by a combination of longitudinal
and transverse ribs. Therefore, the interactions between soil
and geogrid refer to the following mechanisms: (1) shear
resistance between soil and the surface of the geosynthetic
along the reinforcement plane (the ribs, in the case ofgeogrids); and (2) internal shear resistance of the soil (in theaperture of geogrid). The mechanisms (1) and (2) have been
quantified by researchers such as [6], [7].
Practically, the direct shear test is a suitable mean to study theinteraction between soil and reinforcement because it cansimulate the shear mechanism along a potential failure plane
in reinforced earth structure. However, the in-soil interaction
behaviour of a geogrid depends on the relative displacementof the geogrid and soil. It is necessary to use an appropriate
testing method for estimating the behaviour of reinforced soil
structures since the details of the test method may influence
the results of direct shear tests. Moreover, the interaction
mechanisms between soil and geogrid in direct shear mode
are more complicated than sheet-like geosynthetics likemetallic strips and geotextiles. Different test methods havebeen used by different researchers. Mitachi et al. [8] proposeda method to estimate the deformation behaviour of geogrid
embedded in soil based on the results of direct shear tests. The
study also compared the data calculated from direct shear
tests with observed data from pull-out tests. They concluded
that the relative displacement of geogrid in any part of thereinforced soil, as well as the pull-out force vs. pull-out
displacement relationship, can be predicted by using direct
shear test results. Consequently, Nakamura et al. [9]performed a series of direct shear tests changing theconfigurations of soil and geogrid, roughness of substratematerial on which the geogrid is glued, and the distance
between the surface of geogrid and the shear plane. They
proposed that the test specimen of the geogrid glued on thesubstrate should be placed under the sand layer and a suitable
substrate material should be used taking into account themode of interaction between soil and geogrid in situ. A
similar type of device has been used in this study where therelative position of the geosynthetics against the soil layer
remain unchanged which gives the most approximate
condition to that in the field.
The shear strength of sand-geosynthetic interfaces has been
investigated using direct shear tests by a number of
researchers [10]-[12]. These studies used geosynthetic
materials such as High density polyethylene (HDPE), poly
propylene (PP), Polyester (PET) yarns coated with PVC etc.
However, this study focuses on estimating the interface
behaviour of Basalt geosynthetic with sand in direct shear
mode. The frictional behaviour of sand-sand and sand-steel
interfaces is also investigated. Moreover, the interface
efficiency is evaluated with the influence of various
geometric properties of Basalt geogrids at different applied
normal stresses.
2. MATERIALSAs backfill material, air-dry uniform silica sand (Toyourasand), available in Japan, was used in all direct shear tests
conducted throughout this study. Before testing, the physicalproperties of sand were tested in the laboratory. This sand has
no fine content less than 0.075mm with effective diameter D10
of 0.01 mm, D30 of 0.15 mm, D60 of 0.19 mm, the uniformity

19140kPa
120kPa120

80kPa
160kPa

6.0

100

5.0

80

4.0

60

3.0

40

2.0

20

1.0

0

0.0

-20

4.1 Test of sand-sand and sand-steel interfaces
To quantify the shear strength parameters of experimental
sand, the direct shear tests are conducted for four different
normal pressures. The shear stress versus horizontal
displacement and vertical displacement versus horizontal
displacement curves from direct shear tests on sand is plottedin Fig. 4(a). A well defined peak and residual shear stress isobserved at all normal stress level. Maximum shear resistancemobilized within 3-6mm of shear displacement. Before andafter the peak shear stress, hardening and softening behaviour
is markedly observed. The volumetric behaviour shows
positive dilatancy during shear after a small contraction at
initial stage of horizontal displacement. The average peak andresidual angle of internal friction of the sand are found to be
34.5⁰ and 25⁰ respectively. The shear stress-displacement andvolumetric behaviour of sand-steel interface is shown in Fig.
4(b). Several researchers have already studied the interface
behaviour of steel surfaces (rough and smooth) with sand andfine grained soil [15]-[18]. In this study used a smooth steel
plate is used as substrate for holding the geosynthetic
specimens. Thus, it is important to understand the behaviourof smooth steel surface with sand in case of testing geogrid
interfaces, particularly. As shown in Fig. 4(b), small
difference between the peak and post peak shear strength of
the sand-steel interface is observed for all normal stresses.
The average friction angle between sand and smooth steel
surface is calculated as 10.5⁰. All the shear stress lines seemto follow the same line upto yield point which is similar to thefindings reported by Kishida and Uesugi [19] on Toyoura
193

-1.0
0

2

4

6

8

10

12

Shear displacement (mm)

(a)
40kPa
120kPa40

80kPa
160kPa

0.4

30

0.3

20

0.2

10

0.1

0

0.0

-10

-0.1

Vertical displacement (mm)TEST RESULTS

Vertical displacement (mm)

4.2 Test of sand-geosynthetic interfaces
Four types of Basalt geosynthetic materials, including one
woven geotextile (GT) and three geogrids (GG1, GG2 andGG3) have been used to evaluate the behaviour ofsand-geosynthetic interfaces. These interface tests were
performed using the same direct shear equipment which was

Shear stress (kPa)4.

sand-steel interface tests. A progressive nature ofstrain-softening behaviour is also observed after the peak
shear stress. From vertical vs. horizontal displacement
relationship, only compression behaviour is observed during
shearing which is typical of laboratory shearing along smoothinterfaces, e.g. [20].

Shear stress (kPa)plate was used as substrate material on which geosynthetic
specimens were placed below the sand layer. To know thefrictional behaviour of Toyoura sand with the steel plate, thedirect shear tests of sand-steel interface were also performed.
The geosynthetic specimen was positioned on the substrateplaced on the top of the lower box. Subsequently, thespecimen was clamped on the front edge of the steel plateusing four aligned bolts and two steel clamping blocks. The
sand in the upper shear box was rained using a raining device
with two sieves. The dry unit weight of the sand mass inupper box was 1.63 g/cm3 (15.9kN/m3), which corresponds torelative density (Dr) of 95% (dense condition). The thicknessof the sand layer in upper shear box was 70mm. The
specification of this equipment is within the standard of JIS
and the Japanese Geotechnical Society (JGS: T941-199X).
The detail of this equipment is also available at [13]. The
direct shear tests were conducted using normal stresses of 40,
80, 120 and 160 kPa. All shear tests involved applying thenormal stress and monitoring the vertical displacement. The
shear load was only applied after the vertical displacement
had reached equilibrium after applying the normal load. The
vertical load is maintained constant during shearing process.
A shear displacement rate of 0.5mm/min was used during
testing. This rate is in the range recommended in the researchmanual [14]. The maximum shear strength obtained during
the shear process was recorded as the peak shear strength.

-0.2

-20
0

1

2

3

4

5

6

7

8

9 10

Shear displacement (mm)(b)
Fig. 4 Shear stress-displacement and horizontal vs. verticaldisplacement relationships of (a) sand-sand, (b) sand-steel
interfaces

160kPa

0.4

Sand-GT

0.3

40

0.2
20

0.1

0

0.0

-20

-0.1
0

1

2

3

4

5

6

7

80

40

0.4

20

0.2

0

0.0

-20
Shear displacement (mm)
(b)

4.0
3.2
2.4

40

1.6

20

0.8

0

0.0

80
Shear stress (kPa)

60

60

40kPa
80kPa
120kPa160kPaSand-GG3

4.0
3.2
2.4

40

1.6
20

0.8

0

0.0
-0.8

-20

-0.8

-20

-0.2
0 1 2 3 4 5 6 7 8 9 10

Vertical displacement (mm)Shear stress (kPa)

80

0.8
0.6

8

40kPa
80kPa
120kPa160kPaSand-GG2

1.0

60

Shear displacement (mm)
(a)
100

40kPa
80kPa
120kPa160kPaSand-GG1

Vertical displacement (mm)120kPa100

0.5

Vertical displacement (mm)80kPa

Shear stress (kPa)

60

40kPa

Vertical displacement (mm)Shear stress (kPa)80

0 1 2 3 4 5 6 7 8 9 10

0 1 2 3 4 5 6 7 8 9 10

Shear displacement (mm)Shear displacement (mm)

(c)

(d)

Fig. 5 Shear stress-displacement and horizontal vs. vertical displacement
curves for (a) sand-GT, (b) sand-GG1, (c) sand-GG2, (d) sand-GG3 interfaces
used for the direct shear tests on sand and sand-steel
interfaces. The results of sand-geosynthetic interface tests areshown in Fig. 5. It shows that the shear strength of allinterfaces including the geotextile (GT) and geogrids (GG1,
GG2 and GG3) increases with the increase of normal stressesof 40, 80, 120 and 160kPa. At the initial stage of shear
loading, all interfaces develop linear shear stiffness before
yielding takes place. This may be because of the initial highbonding stress between sand and the surface of geosynthetic
materials. From the yield point to maximum shear stress,
strain hardening behaviour is observed in all interfaces. The
post peak nature of sand-GG1 and sand-GG2 interfaces are
similar to that of sand-sand interface which ends up with aconstant residual state. In case of sand-GT interface, a slightstrain-softening response is seen at lower normal stress(40kPa), but this response disappears with the increase of
normal stress and the residual stress becomes equal to thepeak shear stress. Though the sand-GG3 interface shows
strain softening behaviour after the peak, a slight strain

hardening response is seen further for higher normal stressesof 120 and 160kPa. The maximum shear stress obtained fromsand-sand interface tests is consistently higher than thatobtained from all other interface tests except for sand-GG1
interface at lower normal stress of 40kPa. However, themaximum shear stress is found decreasing with the decreaseof geosynthetic area of the geogrids at all normal stresses.
Since GG1 has small apertures and thin ribs, large shear stressis exerted from the initial stage of shear due to interlocking
between geogrid and sand. On the other hand, as the size ofthe apertures of GG2 and GG3 is relatively larger than GG1,
the shear stress is found relatively low due to lessconfinement of sand in the apertures as well as lessgeosynthetic area for effective skin friction.
4.3 Volume change behaviour of the interfaces
The vertical displacement versus shear displacement curves
obtained from the direct shear tests on the sand-geosynthetic
interfaces are shown in Fig. 5. It shows that the maximum
vertical displacement (0.07mm) of sand-GT interface is very194small compare to that of sand-sand interface (0.8mm). The
behaviour is mostly dilative with a little contractive nature at
the initial stage of shearing. The presence of dilation withshearing indicates the presence of some degree of particle
rolling and interlocking as dilation is required for the shearing
and rearrangement of angular particles. The volumetricbehaviour of sand-GG1 interface is seen almost same tosand-sand interface, but the magnitude is different. In case of
GG2 and GG3, as shown in Fig. 5(c) &5(d), the volumetricbehaviour is almost similar to sand-GG1 interface at lower
normal stress. But, exception volumetric behaviour isobserved at high normal stress of 160kPa where GG2 and
GG3 interfaces show very small changes in volume that iscontractive or negative dilatancy in nature. This contractive
behaviour has similarities with that of the sand-steel interface.
It is interesting to note that the maximum verticaldisplacement increases with increasing the percent open areaof the geosynthetics. The maximum vertical displacements
for sand-GT, sand-GG1, sand-GG2 and sand-GG3 interfaces
are 0.07, 0.12, 0.34 and 0.44mm respectively. It means thatthe sand in the apertures of geogrids dominates the dilatancy
behaviour of the sand-geogrid interfaces.

resistance. Although the frictional resistance of sand-GG2
interface is closer to that of sand-GG1 interface, significant
difference is observed for sand-GG3 interface. The possible
reason of this phenomenon is already discussed in theprevious section.
4.5 Interface efficiency
To calculate efficiency of the sand-geosynthetic interfaces,
the peak shear strength of Toyoura sand (τs) at each normal
stress is considered as a reference value. It is to note that theshear strength values are the test values and not the fitted
values using Mohr-Coulomb failure criterion. The shear
strength values of sand-geosynthetic interfaces (τs-g) werenormalized using the reference values at the corresponding
normal stress level (40, 80,120 and 160kPa). The normalized
value is identified as the interface shear strength efficiency orinterface efficiency (Ei= τs-g/τs). In literature similar
relationships have been identified as "bond coefficient" [5],"interface efficiency"[9], and interface shear strength
coefficient [14]. The interface efficiency of different
sand-geosynthetic interfaces are shown in Table 2.
Table 2 Interface efficiency of sand-geosynthetic interfaces

Maximum shear stress (kPa)

4.4 Frictional resistance of the interfaces
The maximum shear stress vs. normal stress values of all six
interfaces are presented in Fig. 6. It can be observed that, if

Interfaces
Sand-sandSand-GG1
Sand-GG2
Sand-GG3
Sand-GT160
Sand‐GT
Sand‐GG2
Sand‐steel

120

Sand‐GG1
Sand‐GG3
Sand‐sand

80
40
0
0

50

100

150

200

Normal stress (kPa)
Fig. 6 Maximum shear stress vs. normal stress for
all interfaces
linear shear strength envelop is drawn fitting the experimentresults it would define a little cohesion intercept. This
intercept may be induced probably due to nonlinearity of theshear strength envelop. The sand-sand interface obtained
maximum friction angle of 34.5⁰ whereas the minimum
friction angle of 10.5⁰ is found for sand-steel interface. The
friction angles of sand-GT, sand-GG3, sand-GG2 andsand-GG1 interfaces are obtained as 18.5⁰, 21.5⁰, 27⁰ and
29.5⁰, respectively. It is observed that among all
sand-geosynthetic interfaces, the sand-GG1 interface
develops the highest frictional resistance whereas sand-GT
interface develops the lowest resistance which is about 86%
and 54%, respectively, of the sand-sand interface frictional

Normal stress, kPa
40kPa
1.00
1.02
0.68
0.59
0.42

80kPa
1.00
0.84
0.85
0.65
0.52

120kPa
1.00
0.81
0.73
0.62
0.49

160kPa
1.00
0.83
0.75
0.56
0.51

Average
1.00
0.87
0.75
0.61
0.49

Table 2 shows that the sand-GT interface strength issignificantly lower than the shear strength of the sand. Unlike
geogrids, the woven geotextile does not allow direct contact
between sand particles. Hence, the shear resistance comes
only from shear between geosynthetic surface and the sandparticles. It is observed that the efficiency of sand-GG1
interface is higher than other sand-geosynthetic interfaces.
Although, at 80kPa of normal stress, the efficiency ofsand-GG2 interface is little higher than that of the sand-GG1
interface. It is interesting to note that the efficiency of GG1
interface at low normal stress (40kPa) is higher (1.02) thansand. But, this is not exception in the literature. Cancelli et al.
[12] reported interface shear strength coefficients ranging
from 1.04 to 1.12 for interfaces between HDPE andpolypropylene (PP) geogrids against sand, Cazzuffi et al. [13]reported interface shear strength coefficient of 0.97 forsand-HDPE geogrid interface, while Liu et al. [21] foundinterface shear strength coefficients for sand-PET (coatedwith PVC) geogrid interface ranging from 0.89 to1.01. For
the sand-geogrid interfaces tested in this study, the averageinterface efficiency ranges from 0.61 to 0.87. However, theinterface efficiency of the same geosynthetic would varydepending on the backfill material used in the field. Thus,
there is scope for further investigation to evaluate theinterface efficiency for different backfill materials likecohesive soils and soil with various admixtures which may
increase interface efficiency.

195

5.

CONCLUSION

This study was carried out to evaluate the interface behaviourof basalt geosynthetic with sand by a series of direct shear
tests. The main conclusions that can be drawn from thisinvestigation are as follows:
1. The interface shear strength of all sand-geosynthetic
interfaces increased with the increasing normal stresses.
2. Strain hardening-softening behaviour is clearly observed
for all sand-geogrids (GG1, GG2 and GG3) interfaces
which resembles the shear stress-displacement behaviourof the sand. The sand-geotextile (GT) interface shows
strain hardening behaviour before peak and almost plastic
condition is observed after the peak.
3. Due to small size apertures and thin ribs of GG1, large
shear stress is exerted from the initial stage of shear. Since
GG2 and GG3 have larger apertures, the shear stress isfound relatively low due to less confinement of sand in theapertures as well as less geosynthetic area for effective
skin friction.
4. The maximum vertical displacement increases withincreasing the percent open area of the geosynthetics
which is 0.07, 0.12, 0.34 and 0.44mm for sand-GT,
sand-GG1, sand-GG2 and sand-GG3, respectively. Itmeans that the presence of sand in the apertures ofgeogrids dominates the dilatancy behaviour of thesand-geogrid interfaces.
5. The interface efficiency of sand-GG1 interface is foundhigher than other sand-geosynthetic interfaces which waseven higher (1.02) than sand itself at low normal stress.
However, the average interface efficiency of thesand-geogrid interfaces tested in this study ranges from
0.61 to 0.87.
6.

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reinforcements in sand”, Geotech. Testing J.,vol.6, No. 3,
1983, pp. 101-111.
[2] Jewell, RA and Milligan GWE,“Interaction between soil
and geogrid”, Polymer grid reinforcement, Conference
publication, Thomas telfold Limited, London, March
1985, pp. 18-30.
[3] Palmeria E M and Milligan GWE, “Scale and otherfactors affecting the results of pull-out tests of grids
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511-524.
[4] Bergado, D. T., Chai, J.C., Abiera, H. O., Alfaro, M.C.,
Balasubramaniam, “Interaction between CohesiveFrictional Soil and Various Grid Reinforcements”,
Geotextiles and Geomembranes, Vol. 12, No. 4, 1993, pp.
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[5] Bergado D. T., Shivashankar R., Alfaro M. C., Chai JC,
Balasubramaniam A. S., “Interaction behaviour of steel
grid reinforcements in a clayey sand”, Géotechnique,
1993, vol.43, No.4, pp. 589-603.
[6] Alfaro M., Miura N., Bergado D. “Soil-Geogrid
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Tests”, Geotech. Test. J., 1995, vol.18, No.2, pp.
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[7] Tatlisoz, N., Edil, TB. and Benson CH., “Interaction
between Reinforcing Geosynthetics and Soil-Tire Chip
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[8] Mitachi, T., Yamamoto, Y. and Muraki, S.,“Estimationof in-soil deformation behavior of geogrid under pull-out
loading”, Proc., Int. Symp. on Earth Reinforcement ,Kyushu Univ., Fukuoka, Japan, vol. 1, 1992, pp.
121–126.
[9] Nakamura T, Mitachi T and Ikeura I, “Direct shear
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[10] Cancelli, A., Rimoldi, P., and Togni, S., “Frictional
characteristi cs of geogrids by means of direct shear andpullout tests”, Proc., Int. Symp. on Earth Reinforcement ,Kyushu Univ., Fukuoka, Japan, 1992, pp. 51–56.
[11] Cazzuffi, D., Picarelli, L., Ricciuti, A., and Rimold, P.,“Laboratory investigations on the shear strength ofgeogrid reinforced soils”, ASTM Spec. Tech. Publ., 1993,
1190, pp. 119–137.
[12] Bakeer, R. M., Sayed, M., Cates, P., and Subramanian,R., “Pullout and shear test on geogrid reinforced
lightweight aggregate”, Geotextiles and Geomembranes,
vol. 16, No. 2, 1998, pp. 119–133.
[13] Hossain, MZ, “Dilatancy behavior of soil-structureinterfaces for farm roads and embankments”, Australian
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pp. 12-17.
[14] Ingold, T.S., “The geotextile and geomembranes
manual”, Elsevier Advanced Technology, ISBN:
1856171981, 1994, pp. 610, U.K.
[15] Kishida, H., and Uesugi, M., “Tests of interface between
sand and steel in the simple shear apparatus”,
Ge´otechnique, vol. 37, No. 1, 1987, pp. 45–52.
[16] Tsubakihara, Y., Kishida, H., and Nishiyama, T.,“Friction between cohesive soils and steel”, Soil andFoundation, vol. 33, No. 2, 1993, pp. 145–156.
[17] Tejchman, J. and Wu, W., “Experimental and numerical
study of sand–steel interfaces”, International Journal for
Numerical and Analytical Methods in Geomechanics, 19,
1995, pp. 513–536.
[18] Uesugi, M., Kishida, H., and Tsubakihara, Y., “Behaviorof sand particles in sand–steel friction”, Soils andFoundations, vol. 28, No. 1, 1988, pp. 107–118.
[19] Uesugi, M., Kishida, H., “Influential factors of friction
between steel and dry sands”, Soils and Foundations, vol.
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[20] Evgin, E. and Fakharian, K., “Effect of stress paths on
the behaviour of sand-steel interfaces”, Canadian
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[21] Liu, C.N., Zornberg, J. G., Tsong C. C., Ho, Y.H. and Lin,
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196

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Effect of Soil Parameters Uncertainty on Seismic Response of Buried
Segmented Pipeline
Chaminda Gallage, School of Urban Development, Queensland University of Technology, Brisbane, Australia
Pathmanathan Rajeev, Department of Civil Engineering, Monash University, Melbourne, Australia
Jayantha Kodikara, Department of Civil Engineering, Monash University, Melbourne, Australia


ABSTRACT: Pipelines are important lifeline facilities spread over alarge area and they generally encounter a range of seismic hazards
and different soil conditions. The seismic response of a buried
segmented pipe depends on various parameters such as the type ofburied pipe material and joints, end-restraint conditions, soil
characteristics, burial depths, and earthquake ground motion, etc.
This study highlights the effect of the variation of geotechnicalproperties of the surrounding soil on seismic response of a buried
pipeline. The variations of the properties of the surrounding soil
along the pipe are described by sampling them from predefinedprobability distribution. The soil-pipe interaction model isdeveloped in OpenSEES. Nonlinear earthquake time-history
analysis is performed to study the effect of soil parameters
variability on the response of pipeline. Based on the results, it isfound that uncertainty in soil parameters may result in significant
response variability of the pipeline.

Keywords: Buried pipe, soil-pipe interaction, soil parameters,
uncertainty, seismic response

1. INTRODUCTION
Buried segmented pipelines are commonly used in many
lifeline systems, such as water distribution and gas supply
systems, etc. There are several publication and reports that
discuss the severe damage to civil life cause by the failure ofburied pipelines during or after the high-intensity earthquake[1], [2]. After the Hyogo-Ken Nanbu earthquake of 1995 in
Japan, it was reported that gas leakage from buried pipelines
occurred at 234 different places; subsequently, fires started
primarily due to gas release and electricity sparks. Fires
occurred at 531 different places and burnt areas were over 1
km2 [3]. Another example, the Chi-Chi earthquake of 1999 in
Taiwan also caused serious damage to natural gas supply
systems. More than 100,000 industrial and residential
customers in the disaster area were cutoff from the natural gas
supply after the earthquake, and the estimated economic lossof five major natural gas companies was approximately
US$ 25 million [4].
During the 1994 Northridge earthquake, several pipelines andaqueducts were broken due to large permanent grounddeformation; and during the 1995 Kobe earthquake, around2000 repairs had to be done in the water distribution system
due to significant ground shaking, ground distortion, andliquefaction in the artificial fills constructed near the bay [5].
Due to the serious consequences of lifeline system failure
under earthquakes, their seismic performance has been thesubject of many research studies. The performance of apipeline under seismic load depends on type of buried pipematerial and joint, end-restraint conditions, soil
characteristics, earthquake ground motion, and burial depths
etc. In this paper, the effect of soil parameters uncertainty inseismic response of pipeline in longitudinal direction isinvestigated. The dynamic soil-pipe interaction is developed
using Winkler-based approach with nonlinear discrete soil
springs in longitudinal and vertical directions. The effect ofsoil variability in seismic response is modeled using
probabilistic approach.
2. PIPELINE CONSIDERATION
A cast iron pipeline of 120 m long with lead caulked pipe
joint buried in sandy soil with fixed end condition isconsidered in this study. The length of pipe segment (ls), yieldstrength (fy), outer diameter (D) and burial depth (H) are 6 m,
250 MPa, 150 mm and 800 mm, respectively. The friction
coefficient (δ), depending on the outer-surface characteristics
and hardness of the pipe, is taken as 0.8. It is assumed that thepipeline is placed well above the ground water level;
therefore, the soil liquefaction is not considered. The change
in internal pressure in the pipeline and the live load over the
ground surface are neglected.
3. FINITE ELEMENT MODELING OF PIPELINE
Finite element analysis of the pipe-soil system is performed
using OpenSEES finite element analysis package [6]. Pipesegment is modeled with elastic beam element and the joint ismodeled using zero-length element with nonlinear material
model. Also, the soil response in vertical and axial direction ismodeled with zero-length element and nonlinear material
model. The schematic diagram of pipe-soil system is shownin Fig.1.

197

Fig.1. Schematic diagram of pipe-soil system

3.1 Behavioral model for soil spring
The force-displacement behavior of axial soil springs are
based upon recommendations in the 1984 ASCE Guideline
for Gas and Liquid Fuel Pipelines [7] for non-cohesive
backfill. The nonlinear behavior of axial spring is shown in
Fig. 2 (right).
A symmetric bilinear-type curve (elastic-perfectly plastic) is
employed with a peak force per unit length at the soil-pipe
interface, Tu:

Tu 


2

DH (1  K0 ) tan( )N  exp(0.18  2.5)

(6)

The displacement, Δd , at Qd is 0.1D for granular soil. A moredetailed description of (1) to (6) is provided in the American
Lifeline Alliance, ASCE [8].
3.2 Behavioral model for pipe joint
The axial force-deformation relationship considered for thelead caulked pipe joint is based on the study reported in [9].
The force-displacement response is represented by bi-linear
relationship as shown in Fig. 3.

(1)

where γ is the effective unit weight of the soil,  is the angle
of internal friction of the sand and Ko is the coefficient oflateral soil pressure at rest. The equivalent “yield”
displacement (Δt) for the soil spring is typically about 2 mm.

Fig.3. Axial force-displacement behavior of joint springFig.2. Nonlinear force-displacement behavior of soil springs:
vertical spring (left); axial spring (right)
The nonlinear behavior of vertical soil springs is shown in Fig.
2 (left). The upward soil resistance per unit length (Qu) of thepipe in sandy soil can be determined as:Qu  N qvHD
N qv 

H
44D

(2)

 Nq

(3)



N q  exp( tan  ) tan 2  450  
2


(4)

The corresponding displacement, Δu at Qu can be taken as
0.01H for the dense sand.
The vertical bearing soil spring force per unit length (Qd) can
be calculated by:

Qd  N q γHD  N γ γ

D2
2

(5)

There is an initial elastic region until slippage at a relative
joint displacement of Δs, followed by a linear post-slippage
region until leakage at a relative joint displacement of Δl. The
axial force at slippage, Fs, is based upon a model proposed in[10].

Fs  CaDd1(7)

where Ca is the adhesive strength at the pipe/lead interface
and d1 is the depth of lead caulking. The average value of Ca
is around 1.7 MPa, and d1 for a 150 mm diameter pipe is 55
mm. The initial stiffness of the joint and the joint force atleakage also reported in [9]. For 150 mm diameter pipe theinitial stiffness is quite large, resulting in a typical slippage
displacement of only about 3.3 μm. The joint force at leakage
is equal to twice the slippage force. The slippage
displacement at leakages (Δl) is taken as 16.5 mm.
4. EARTHQUAKE GROUND MOTION
In order to perform nonlinear time-history analyses, theacceleration time history recorded at El-Centro during
Imperial Valley California (1940) is considered. The
characteristics of the record are: the magnitude (Mw) is 6.6,
source-to-site distances (r) range is 8 km and soil condition isstiff. The maximum acceleration to velocity ratio (A/V) is
1.04, where A is in 'g' and V is in m/s. Fig.4 shows theacceleration time history and acceleration response spectrawith 5% damping of El-Centro record.

198

Fig.5. Histogram of sampled soil unit weight (left) andfraction angle (right)

Fig.4. El-Centro acceleration time history (top) and response
spectrum (bottom)

Each randomly generated (γ,φ) pair is input into the finite
element model and time history analysis is performed to
predict the response. The natural period of the pipeline withmean value of soil property is 0.27 s.

5. VARIABILITY IN SOIL PROPERTIES AND THEIRMODELING

7. RESULTS AND DISCUSSION

It is evidence from section 3.1, that the soil proprieties are oneof the performance controlling parameter of pipe-soil system.
The force-displacement behavior of soil spring is controlled
by the soil parameters. Thus, it is essential to investigate theeffect of soil parameters uncertainty in seismic response ofburied pipelines. In this study, the soil properties such as
internal friction angle and unit weight of soil are considered
as uncertain variable. Those variables are assumed bedescribed by lognormal distribution, since they arenecessarily positive values. Table 1 shows the list andrespective mean values and corresponding 10% coefficient ofvariation (CV) of the input parameters. Although somestudies have reported that a larger variation of the in situ CV,
the 10% CV for different soil types including, sand, clay andsilt seems to be reasonable [11]. No correlation between
internal fraction angle and unit weight is considered.

Fig.6 shows the axial displacement of pipeline, which has themean values of the soil properties, at different time in the time
history analysis. The displacement is concentrated at thejoints and the maximum displacement is occurred at the
middle section of the pipeline. The maximum pipe
displacement observed at the middle of the pipeline isapproximately 24 mm.

Table.1 Statistical property of soil parameters
Parameter
Mean
CV (%)
Distribution
Friction angle
350
10
Lognormal
Unit weight
16 (kN/m3)
10
Lognormal
6. TREATMENT OF UNCERTANITY IN SEISMICRESPONSE ANALYSIS
Seismic response analysis of the pipeline which incorporates
the source of uncertainty considered above can be carried out
using random sampling Monte Carlo simulation coupled withthe finite element pipeline model. Fig. 5 shows histogram of
500 values of random sampled fraction angle and unit weightof soil.

Fig.6. Axial displacement of pipeline at different time in thetime history analysis
The maximum relative displacement is observed at the joints
which are close to the supports. Fig.7 shows the maximum
relative displacement at each joint along the pipeline. The
maximum relative displacement at the joints is close to 9 mm
which is less than the slippage displacement at leakages. The
joint at the middle of the pipeline shows no relative
displacement for the mean valued model. The maximum
relative displacement is high at the joints close to support anddecreases towards the middle of the pipeline.

199

section of the pipeline show no relative displacement forupper value of fraction angle. Contrary, the pipeline shows
significant increase in relative joint displacement for lower
value of frication angle from the mean valued model. Further,
the effect of fraction angle on the response of the system ismuch significant compare to the unit weight of the soil.

Fig.7. Maximum relative displacement at the joint during
time history analysis
To study the effect of single soil parameter on the response,
the lower and the upper value among the sampled values ofeach parameter is selected, while keeping the other one in its
mean. Fig. 8 shows the effect of unit weight of soil on the
response. For the upper value of the unit weight, themaximum relative joint displacement response is increased ingeneral compare to the mean valued model response anddecrease for lower value of the unit weight. It may be due tothe fact that the mass of the system is increased withincreasing unit weight and consequently the system subjected
to higher inertia force, which causes larger displacement in
the system. In the case of lower value of unit weight of soil, a
few joints at the middle part of the pipeline show no relative
displacement.

Fig.9. Effect of friction angle on the maximum relative
displacement at the joint
Fig. 10 shows the 95% confidence bound and median of the
maximum relative joint displacement response from therandom sampled model together with the mean-valued model
response. The median and mean-valued model responses are
essentially the same. But the variability in the response due
the uncertainty in the soil properties is significantly high.

Fig.10. Mean-valued pipeline model response and 95%
confidence bound and median of random sampling model
response

Fig.8. Effect of unit weight on the maximum relative
displacement at the joint
Fig.9 shows the effect of internal fraction angle of soil on the
maximum relative joint displacement response. For the lower
value of the soil fraction angle, the joint displacement isincreased and for the upper value of the fraction angel the
joint displacement is decreased significantly from the meanvalued model. Due to large axial soil fractional forces in thepipe segment, a few joints at each side from the middle

8. CONCLUSION
The influence of soil parameters uncertainty on the seismic
response of buried segmented pipeline is investigated in thispaper. A 120 m long pipeline in sandy soil is molded using
nonlinear Winkler approach to represent the soil behavior.
Random sampling method is used to randomize the soil

200

parameters. The effect of different types of soil is notconsidered. Within this limitation, the analysis results showthat the uncertainty in soil property has significant influenceon the displacement response of pipeline. The influence ofinternal fraction angle is significant compare to unit weight ofthe soil. The coefficient of variation of joint response is morethan 100% in some location. Finally, it is important toconsider the variability in soil properties in seismic risk
assessment of pipeline.
9. REFERENCES[1]

Ayala G and O’Rourke MJ, “Effects of the 1985 Michoacan earthquakeon water systems and other buried lifelines in Mexico”, Report No.
NCEER-89-0009, National Center for Earthquake Engineering
Research, State University of New York at Buffalo, 1989.
[2] Wang L, Shao-ping S and Shijie S, “Seismic damage behavior of buried
lifeline systems during recent severe earthquakes in U.S., China andother countries”, Technical Report No. ODU Lee-02, Old Dominion
University Research Foundation, 1985.
[3] Scawthorn C and Yanev PI, “Preliminary report on Hyogo-ken Nambu,
Japanese earthquake” Eng Struct, Vol. 17(3), 1995, pp.146-57.
[4] Chen WW, Shih BJ, Wu CW, “Chen YC. Natural gas pipeline systemdamages in the Ji-Ji earthquake (The City of Nantou)”, In: Proc of the
6th international conf on seismic zonation, 2000, pp ..
[5] Eidinger JM. and Avila EA, Guidelines for the Seismic Evaluation andUpgrade of Water Transmission Facilities. Monograph # 15. TechnicalCouncil on Lifeline Earthquake Engineering, American Society of
Civil Engineers, 1999.
[6] McKenna F, Fenves GL, Jeremic B and Scott MH,Open System for
Earthquake Engineering Simulation, http://opensees.berkely.edu, 2007.
[7] American Society of Civil Engineers, Guidelines for the Seismic
Design of Oil and Gas Pipeline Systems Reston, VA, 1984, pp. 473..
[8] American lifelines alliance, Guidelines for the design of buried steel
pipe. ASCE, 2001.
[9] Hmadi EI and O’Rourke, MJ, Seismic Wave Effects on Straight Jointed
Buried Pipeline. Technical Report NCEER-89-0022, Multidisciplinary
Center for Earthquake Engineering Research, New York,1989[10] O’Rourke TD and Trautmann CH, Analytical Modeling of Buried
Pipeline response to Permanent Earthquake Displacements, Report No
08-4, School of Civil Engineering and Environmental Engineering,
Cornell University, Ithaca, New York, 1980.
[11] Jones AL Kramer SL and Arduino P, Estimation of uncertainty ingeotechnical properties for performance-based earthquake engineering.
Technical Report 2002/16, Pacific Earthquake Engineering ResearchCenter, PEER, 2002.

201

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Sorption Capacity of Cu by SC-Soil from Batch and Column Tests
Arwut Yimtae, Pongsakorn Punrattanasin and Noppadol Sangiumsak
Faculty of Engineering, Khon Kaen University, Thailand

ABSTRACT: This paper presents the sorption capacity of
Cu by SC-soil collected from a location in the Northeasternpart of Thailand. Copper (Cu) is one of the common heavy
metals that have been discharged from industries and fromthe use in the daily life of human being. The toxic
conditions from Cu are always occurred with plants,
animals, and human from this metal. The batch equilibrium
method was used as a tool to investigate the adsorption
value. The results from batch tests can demonstrate thesorption isotherm of the soil. The results showed that theadsorption isotherm well fit both the Langmuir andFreundlich models. The next series of test were evaluatedby performing column leaching test. The purpose of the test
is to investigate the transportation phenomena of liquid into
soil layer. This method is also for the assessment of groundwater risk due to pollutant emission from contaminatedsoil.

surface of soil particles are almost rough. Table 2
summarized the compound composition of this soil byusing XRF technique.

Fig.1 The aerial view of the location of excavation pit

Keywords: Copper, Adsorption, Isotherm, Langmuir
1. INTRODUCTION
Due to economic growth during many decades in Thailand
especially in the Northeastern part, industrial sector leads
many environmental problems. Industries discharged thewaste water into river or soil without treatment and care.
Copper (Cu) is one kind of heavy metals and toxins that hascaused serious health problem to humans and animals [1,2].
There are many methods for remove the heavy metals fromthe waste water and soil [3,4,5]. Adsorption is one of themost common methods for heavy metal removal because itis an effective and economical technique. This is the main
objective that brought about the research to be a choice ofimproving pollution problem by focusing on the sorption
capacity of heavy metal by soil. In this work, the adsorptionof Cu by soil found in the Northeastern part of Thailand wasstudied and investigated by batch and column techniques.

Fig.2 The collection of soil sample

2. MATERIALS AND METHODS
2.1 Adsorbent
The material used as adsorbent was soil collected from one
location in the upper part of Northeastern (Fig.1), Thailand.
It was excavated at a depth of 50 cm from its surface (Fig.2).
The soil was dried at a temperature of 110 oC for 48 hours
and was sieved through sieve No.16. It can be classified by
the Unified Soil Classification System (USCS) as SC-Soil.
It will then be called SC-Soil throughout this paper.
Engineering properties and porous properties of the soil
sample are tabulated and shown in Table 1. The porous
properties were evaluated by adsorption of nitrogen gas(N2) at -196 oC. The specific surface area was calculated by
Brunauer-Emmett-Teller (BET) equation. Figure 3 shows
the SEM photographs of the soil with different
magnifications. As shown in Fig 3, it can be seen that theTable 1. Engineering properties of soil

203

Soil
Specific gravity
Optimum moisture content(OMC), %Maximum dry unit weight (  d ),kN/m3
Coefficient of permeability (k),cm/sSpecific surface area (SBET), m2/g
Micropore volume (VDR), cm3/g
Total pore volume (VT), cm3/g
Average pore diameter (DP), nm

SC-Soil
2.65
18.28
16.48
0.51 x10-6
103
0.040
0.21
8

(a) x300(e) x5000
Fig.3 SEM photograph of SC-Soil at different
magnifications

Table 2. Compound composition of SC-soil(%)(b) x500

O
51.26

Si
36.89

Al
8.62

K
1.54

Ti
0.61

Ca
0.39

P
0.37

Mg
0.31

2.2 Heavy Metal Solution
Heavy metal chosen as the adsorbed materials in this study
was copper (Cu). It was prepared by dissolving Copper
Nitrate (Cu(No3)2) in distilled water. The solutions werevaried with concentrations of 5, 25, 50, 100, 250 and 500
mg/L. Table 3 summarized properties of copper nitrate.
Selection of Cu as a heavy metal was based on its low
toxicity and also Cu is easily found in all areas and Cu is
commonly discharged from various industries.
Table 3. Properties of copper nitrate
CompoundFormulaMolecular weight (g/mol)
Density (g/cm3)
Solubility (g/100 ml)(c) x1000(d) x3000

Copper NitrateCu(NO3)2
295.652.07
137.8

2.3 Experimental work
In this study, batch and column adsorption tests werecarried out to investigate the heavy metal adsorption. The
batch test processes can be performed by mixing a 2.5 g ofsoil with the Cu solution. The initial concentrations wereranged from 5 to 500 mg/L and put it in 120 mL plastic
bottle. Then the mixture was shaken by the horizontal
shaker with a speed of 130 cycles per minute for 0.5, 1, 3, 6,
12, 24, 48, and 72 hours. After a particular period of thetime, the soil was separated from the heavy metal solution
by using a 0.45µm filter and diluted it into the solution.
Finally, concentration of the residual solution was analyzed
by using Atomic Adsorption Spectrometer (AAS) (Fig.4).
From the step above, it can be measured the adsorption
equilibrium time and the amount of heavy metal adsorbed
by soil at equilibrium time (q). The equilibrium amount ofheavy metal can be calculated from the following equation:

Fig.3 SEM photograph of SC-Soil at different
magnifications
204

q

(Co  Ceq )Vsol

(1)

Ms

Where q is the adsorption capacity of the soil at equilibrium
condition (mg/g), Co is the initial concentration of heavy
metal solution (mg/g), Ceq is the equilibrium concentrationof the solution (mg/g), Vsol is the volume of solution (cm3),and W is the mass of soil (g). After the adsorption volume
was determined, the adsorption isotherm can then be
investigated. The adsorption isotherm is the relationshipbetween the concentration of the heavy metal solution atequilibrium condition (Ceq) and the equilibrium amount ofheavy metal adsorbed by soil (q). In this study, theLangmuir and Freundlich models were used to describe theadsorption isotherm.

solutions leaked under the sample were then collected andwere sent to the lab to measure the concentration by AAS
technique. In the second cycle, the distilled water was used
as a solution instead of copper. The purpose of this cycle is
to simulate the leaching effect from raining. The adsorptionof copper around the soil surface was leached by this water.
The remaining concentration can then be measured. The
results of this effect are depending on the soil type and its
physical and chemical properties. The time used in thiscycle is the same as in the first cycle. The last cycle wasthen performed by using the copper solution again. The
purpose of the last cycle is to simulate the repetition ofleaking of contaminated solutions into the soil. The
amounts of leaking solution from the bottom of thecontainer were then sent to measure the concentration after
a particular period of times.

Fig.4 Atomic Adsorption Spectrometer (AAS)
For column leaching test, the purpose of the test is toinvestigate the transportation phenomena of liquid into soil
layer. The column test is commonly used for thedetermination of desorption or dissolution rates ofcontaminants from contaminated soil. This method is also
for the assessment of ground water risk due to pollutant
emission from contaminated soil. The column leachingprocedure used in this study is described briefly in Fig.5.
Figure 5 show the schematic drawing of equipment used in
the column test. The soil sample was compacted bystandard compaction method into the column with amaximum dry density and optimum moisture content.
Contaminated water is pumped and put in the top container.
The soil column and storage tank are connected each otherby system of nut. The flow velocity of contaminatedsolution and water in the column are adjusted by pressure
from pressure pump. In this study flow velocity in thecolumn was about 1 m/day, which corresponds roughly totypical groundwater flow velocity. The advantage ofcolumn test compared to the batch technique is that theamount of soil used in the column is much larger than in thebatch test and the column technique can simulate theconditions as found in the real field.
In this study, three cycles of column test were done tosimulate the infiltration of contaminated solution into thesoil. In the first cycle, copper solution was used and thesolution was infiltrated into the soil under pressure. The
concentrations were measured with different elapsed times.
In this cycle, a total time of 144 hours or 6 days were used.
The elapsed times of 1, 3, 6, 12, 24, 48, 96 and 144 hours
were selected to pick up the sample concentrations. The
205

Fig.5 The schematic drawing of column test

3. RESULT AND DISCUSSION150
y = 0.3034x + 7.4986
R² = 0.9636

3.1 Adsorption isotherm from batch test
The adsorption of Cu for various concentrations by SC-Soilat different time was shown in Fig 6. The results show theremaining concentration of the heavy metal (Ct) as afunction of time. It can be seen that the adsorption took
place rapidly at the beginning of the reaction, which theconcentration rapidly decreased at the period of 1-3 hours
and decline over time until reaching the equilibrium
condition within 6-12 hours.

Ceq /q (g/l)

50

0

500

5 mg/L50 mg/L
250 mg/L

400

Ct (mg/L)

100

025 mg/L
100 mg/L
500 mg/L

100

200

300

400

Ceq (mg/L)
Fig.8 Langmuir adsorption isotherm of Cu

300

10

200
100

y = 0.4048x0.3563
R² = 0.9948

q (mg/g)

1

0
0

12

24

36

48

60

72

84

0.01

0.1

Time (hr)

1

10

100

1000

0.1

Fig.6 Effect of contact time on Cu adsorption by SC-Soil
Figure 7 showed the adsorption isotherm of Cu. As shownin 6, the amount of adsorption (q) increased with increasing
equilibrium concentration (Ceq). The adsorption isotherm
can be represented by Langmuir and Freundlich adsorption
isotherms as shown in Figs 8 and 9, respectively. The
consistency of the isotherm describes that the surface ofsoils was covered with monolayer of the metal particles.
Langmuir parameters ( and ) and the coefficient ofcorrelation (R2) were then summarized in Table 4.
Freundlich parameters were summarized in Table 5. Veryhigh coefficient of correlation (R2) from both models
indicated that the adsorption of Cu by soil sample favorable
fit both to Langmuir and Freundlich adsorption isotherms.

0.01

Ceq (mg/L)
Fig.9 Freundlich adsorption isotherm of Cu
Table 4. Langmuir parameters of soil samples
SoilSC

 (mg/g)
3.2960 (L/mg)0.0405R2
0.9636

Table 5. Freundlich parameters of soil samples

4

Soil
SC

K (L/g)0.4048n2.8066R2
0.9948q (mg/g)

3
2
1
0
0

100

200

300

Ceq (mg/L)
Fig.7 Adsorption isotherm of Cu adsorption by SC-Soil400

3.2 The results from column test
The relationship between sorption capacity of Cu and flowrate of polluted solution with a particular time (Ct) are
shown in Figs. 10 and 11, respectively. At the first cycle,
the initial concentration of 100 mg/l was used. In Fig.10, it
can be seen that the measured values of remaining
concentration, Ct at a time of 0 hour is approximately 0.9
mg/l. It means that the soil has a high potential to adsorb
almost all of the concentration. After a certain period oftime, the measured concentrations seem to slightly andslowly increase with time. The reason to explain thisphenomenon is that the accumulations of Cu particles on
the soil surfaces almost reach the maximum value that soil
can offer so that the areas for the adsorption remain smaller.
The rate of adsorption found is very slow (1.6 mg/l). In the206second cycle (144-288 hours), the distilled water was
replaced in the container. The purpose of this cycle is tosimulate like a raining occurred above the soil surface. It
can be seen in the middle part of Fig.11 that theconcentrations of the solution are almost the same as at theend of the first cycle but Ct is not 0. It is indicated that Cu
still slightly leaked by the leaching effect of water. In thelast cycle (288-432 hours), the copper solution was put into
the container again. The measured values of concentration
seem slightly increase with time and have a same trend
found in the first cycle. It can be said that after leaching bywater, the soil can still adsorb copper and the soil did not
lost the capacity in adsorbing the copper solution. In Fig. 12,
it can be found that the flow rates are in the range of 4-7
cm3/hr.
2.5

5 ACKNOWLEDGMENT
This research was funded by Faculty of Engineering, Khon
Kaen University. The authors wish to acknowledge andgrateful thank to Dr.Arthit Neramittagapong and
Ms.Wimomporn Iamamornphonth of the Department ofChemical Engineering, Khon Kaen University.

2
1.5

Ct (mg.L)

can favorable and well describe the adsorption of this soil.
For column leaching test, the purpose of the test is toinvestigate the transportation phenomena of liquid into soil
layer. In this study, three cycles of column test were done tosimulate the infiltration of contaminated solution into thesoil. In the first cycle, copper solution was used and thesolution was infiltrated into the soil under pressure. The
concentrations were measured with different elapsed times.
In the second cycle, the distilled water was used as asolution instead of copper. The purpose of this cycle is tosimulate the leaching effect from raining. The adsorption ofcopper around the soil surface was leached by this water.
The remaining concentration can then be measured. The
last cycle was then performed by using the copper solution
again. The purpose of the last cycle is to simulate therepetition of leaking of contaminated solutions into the soil.

1

0.5
0

6

REFERENCES[1]

Mohumed EI Zayat and Edward Smith, “Removal of heavy metals
by using activated carbon produced from cotton stalks,” Canadian
Journal on Environmental, Construction and Civil Engineering, vol.
1, September. 2010, pp. 71-79.
Y. B. Onundi, A. A. Mamun, M. F. Al Khatib and Y. M. Ahmed ,“Adsorption of copper, nickel and lead ions from synthetic
semiconductor industrial wastewater by palm shell activated carbon,”International Journal of Environmental Science and Technology,
vol. 7(4), 2010, pp. 751-758.
Souag R, Touaibia D, Benayada B and Boucenna A, “Adsorption ofheavy metals (Cd, Zn and Pb) from water using keratin powder
prepared from Algerien sheep hoofs,” European
Journalof
Scientific Research, vol. 35, No.3. 2009, pp. 416-425.
Kailas L. Wasewar, “Adsorption of Metals onto Tea Factory Waste:
A Review,”IJRRAS, vol. 3(3), June 2010, pp. 303-322.
Tae-Young Kim, Sun-Kyu Park, Sung-Yong Cho, Hwan-Beom Kim,
Yong Kang, Sang-Done Kim and Seung-Jai Kim, “Adsorption ofheavy metals by brewery biomass,” Korean Journal, Chemical
Engineerin, vol. 22(1). 2005, pp. 91-98.

[2]

0

144

Time (hr)

288

432

Fig.10 The sorption of Cu at different times by column test

[3]

8

[4]

7

[5]

q (cm3/hr)

6
5
4
3
2
1
0
0

144

Time (hr)

288

432

Fig.11 The flow rate of Cu at different times by column test

4

CONCLUSION

In this study, the adsorption capacity of Cu by SC- Soilcollected the upper part of Northeastern, Thailand wasinvestigated by performing batch and column leaching tests.
The results from batch test indicated that the adsorption
reaches the equilibrium time within 6-12 hours. The
adsorption capacity increased with increasing initialconcentration. Langmuir and Freundlich isotherm models
207Construction Materials

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Green Construction: The Role of Palm Oil Fuel Ash in Concrete
A.S.M. Abdul Awal, Department of Structure and Materials, Faculty of Civil Engineering
Universiti Teknologi Malaysia, 81300 Skudai, Johor, Malaysia

ABSTRACT: The pressing need for the preservation of natural
resources and reductions in carbon dioxide emissions together withthe ever-rising production volume of concrete to meet the needs ofthe developing world have fuelled the search for alternative
solutions. As the green construction movement begins to hit its stride,
recycling of waste and utilization of supplementary cementing
materials are gaining promises as efficient and attractive choices in
the production of concrete. This paper is a state-of-the-art report on
the study and research findings of the behaviour of concrete
containing palm oil fuel ash (POFA) as supplementary cementing
material. Fresh and hardened state properties including durability
aspects of POFA concrete in both indoor and outdoor exposure
conditions have been presented and discussed. Along with thetypical applications, possible use and potential benefits of the highvolume palm oil fuel ash in concrete have also been highlighted inthis paper.
Keywords: Supplementary cementing materials,
development, palm oil fuel ash, green construction

sustainable

1. INTRODUCTION
Over the decades there has been a growing emphasis on theutilization of waste materials and by-products in constructionindustry. Use of these materials not only helps in getting themutilized in cement, concrete, and other construction materials,
but also has numerous indirect benefits such as reduction inland-fill cost, saving in energy and protecting theenvironment from possible pollution effects. Further, theirutilization has shown to improve the microstructure,mechanical and durability properties of mortar and concrete,
which are difficult to achieve by the traditional method of
concrete construction.
There is no doubt that the inclusion of pozzolanic materials,
either naturally accruing or artificially made, as a partial
replacement of cement imparts significant enhancements ofthe basic characteristics of the resulting mass both in its freshand hardened states. Of all the silicon by-products, pulverizedfuel ash or fly ash is widely used globally. Added to this is thefact that fly ash makes significant changes to physical,
mechanical and durability properties of concrete that are well
documented in national codes and standards. Apart fromindustrial waste, ashes from agricultural origin like rice husk,
coconut husk, corn cob, peanut shell etc. have been identified
as supplementary cementing materials in many parts of theworld [1]-[4]. One of such pozzolanic materials is palm oil
fuel ash, a waste material obtained on burning of palm oil
husk and palm kernel shell in palm oil mill plants.
The objective of this paper is to provide a cohesive literature
survey on studies published in recent decades, accounting forthe contribution or behaviour of palm oil fuel ash (POFA) in
concrete, rather than a comprehensive report summary of all
the investigations. Being a new member in the ash family,

however, POFA has not been studied in all types of concrete
works particularly at long-term outdoor exposure conditions.
Considering the availability and the inherent quality,
significant research findings have been highlighted in regard
to the utilization of this ash as a sustainable construction
material in concrete.
2.

CONCRETE AND ENVIRONMENT

2.1 Waste Materials in Concrete Construction
Concrete is without question one of the oldest and strongest
man-made materials in the world where cement as a binding
material plays a key role in achieving strength and durability.
Portland cement concrete is a fundamental and versatilebuilding material that touches the life of nearly every
individual in the modern world. It is difficult indeed tovisualize what a comparatively dark age we would be living
in without concrete. The economy of many nations ismarkedly influenced by the cement and concrete industry. It
has been estimated that the world-wide consumption of
concrete amounts to about one thousand kilogram per person,
and this demand is expected to rise in future [5]. However theenvironmental aspects involved in the manufacture and use of
cement, concrete and other building materials are of growing
concern. This situation can, however, be improved byeffective and efficient utilization of existing materials
together with the development of recycled materials
generated as industrial or agricultural wastes.
With a given set of constraining conditions waste materials inconcrete mixes can be employed in several ways. For many
purposes a pozzolan, for example, has been regarded as a
substitute for a proportion of cement in a concrete. In somecases, it is preferable to use it as an addition. This involves
direct weight addition of pozzolan to cement, replacing partof aggregate (usually sand) in concrete. Another method
which Samarin et al. [6] describe as a replacement-addition
method that involves replacement of a part of the Portland
cement with excess weight of fly ash, replacing also part of
the aggregate in order to achieve more economical concrete.
According to Dunstan [7], fly ash should be considered to be
the ‘fourth ingredient’ in concrete, that is in addition to theaggregate, cement and water but not as a replacement of thecement. It has also been postulated that only in this way it ispossible to obtain the fullest benefits from fly ash,particularly with low lime content. Whatever is the mode ofapplication, all the methods can result in a significant
improvement and optimization of certain properties of both
fresh and hardened concrete. The strategy of applying
pozzolans in concrete, as revised and modified from Hwang[8], is sketched in Fig. 1.

211

 Reduce
water-cement
ratio
 Low heat ofhydration
 Moreeconomical

PartialReplacement of
Cement Increasewater-cement
ratio
 High earlystrength Moreeconomical

PartialReplacement ofSand

POZZOLAN

PartialReplacement ofboth Cement
and Sand Low heat of
Hydration
 High earlystrength Moreeconomical

Addition of
Pozzolan Reducewater-binder
ratio with the aidof superplasticiser
 High early andultimate strength Additional cost

Fig. 1 Strategies of using pozzolans in concrete productions
Although the use of supplementary cementing materials has
been increased during the last decades, the amount is still
below the satisfactory level. At the same time the data
available so far on the performance behaviour of suchmaterials are not ample; the current information cansometimes be confusing regarding their usefulness. In
response to growing environmental economic forces it is,therefore, likely that the materials that might have been
researched to some extent will require adaptation and further
development for full commercial utilization. In this respect,green building practices have all the potentials towards
sustainable construction.
2.2 Green Construction and Sustainable Development
There is no single definition of green construction. But, ingeneral green construction or green building refers to a
structure using process that is environmentally responsible
and resource-efficient throughout a building’s life-cycle:
from sitting to design, construction, operation, maintenance,
renovation and demolition. Green building aims to maximize

efficiency in their use of water, energy and other resources tominimize waste, pollution or other contributions toenvironmental degradation, and to create environments that
contribute to health and productivity.
Although the concept of green construction differs from place
to place there are fundamental principles that consist of
economy, utility, durability and comfort [9],[10]. The essenceof green building is thus an optimization of one or more ofthese principles. As the world population continues to grow,further urbanisation and developments are inevitable. The
challenges in civil engineering practices in the near future
will, therefore, be to realise projects in harmony with theconcept of sustainable development, and this involves the useof high performance materials and products manufactured
at reasonable cost with the lowest possible environmental
impact. Concrete is, indeed, the most widely used man-madematerial in the world, and is second only to water as the mostutilised substance on the planet. However, the production ofPortland cement, an essential constituent of concrete, releases
large amount of CO2 into the atmosphere, i.e. about one ton of
CO2 for every ton of Portland cement produced. As CO2 is amajor contributor to the green house effect and the global
warming of the planet, both the developed and developing
countries are considering very severe regulations andlimitations on CO2 emissions.
To make a breakthrough the trend toward sustainable
construction has pushed the envelope in the area of concrete
mix designs by introducing materials that are not only more
green but also can create a more superior final product. There
are several factors that constitute ‘green’ and the overall
greenness of a new construction can be increased
exponentially through synergic actions of innovative
construction materials. The first and simplest measure insustainable construction is the choice of materials with highpercentage of recycled content. As a normal practice fly ash,for example, is used to partially replace Portland cement at
relatively modest levels of 15 to 25% (by mass). Research
workers have demonstrated that fly ashes can suitably be used
at much higher replacement levels of between 40 to 60% toproduce concrete with good mechanical properties andexcellent durability [11]-[14].
2.3 Palm Oil Fuel Ash: A Unique Cement Substitute
The oil palm is a tall-stemmed tree which belongs to palmfamily Palmea. The countries in the equatorial belt thatcultivate oil palm are Benin Republic, Colombia, Ecuador,
Nigeria, Zaire, Malaysia, Thailand and Indonesia of which
Malaysia is the largest producer of palm oil and palm oil
products. To date, for instance, there are more than two
hundred palm oil mill plants operating in the country that are
self sufficient industry as far as energy utilisation is concern.
It has been estimated that the total solid waste generated bythis industry has amounted to about ten million tons a year[15]. The palm fibre and shell obtained as waste products by
the industry are generally used as boiler fuel to produce steam
for electricity generation and palm extraction process. The
ash produced on burning palm fibre and palm kernel shell is
considered to be a waste product, the disposal of which poses
enormous problems. As a normal practice this ash, popularly212

known as palm oil fuel ash (POFA) is simply disposed ofwithout any commercial return. However, it has been
identified that POFA has good pozzolanic properties that canbe used as a cement substitute in mortar and concrete mixes[1],[3],[16].
In bulk, POFA is grayish in colour that becomes darker withincreasing proportions of unburned carbon. The particles
have a wide range of sizes, a typical electron micrograph of
POFA being shown in Fig. 2. Physical properties and thechemical analysis of typical ash, illustrated in Table 1, revealthat POFA satisfies the requirements to be pozzolanic andmay be classified in between Class C and Class F according to
the standard specified in ASTM C6I8-94a. Critical evaluation
based on the current literature, however, suggests that thepresent-day classification system of grouping of ashes into
Class N, Class F, and Class C is not adequate to apprise their
total usefulness. It should be the performance-oriented
criteria rather than the chemical composition alone should beconsidered for proper representation of ashes of agricultural
origin like POFA.

3. FRESH CONCRETE PROPERTIES
3.1 Setting Time of OPC and Palm Oil Fuel Ash
One of the most important properties of cementing materials
is setting time that influences the fresh and hardened state
properties of concrete. The setting time of OPC as well as of
OPC with POFA at various replacement levels are shown in
Table 2. Setting times given in the table suggest that like otherfly ashes the inclusion of POFA in cement paste retards thesetting time; the higher the amount of ash the more is theretarding time. This is to be expected; because when the ash
content is high, less is the amount of cement that results inslower setting time. Test results on setting time also revealthat setting time of cement with 60% ash is within the limit as
specified in ASTM C150-94. The slight retarding effect withhigher amount of ash would, however, be beneficial toconcreting in hot weather conditions.
Table 2: Setting time of OPC and palm oil fuel ash
Mix
100% OPC
90% OPC + 10% POFA
80% OPC + 20% POFA
70% OPC + 30% POFA
60% OPC + 40% POFA
50% OPC + 50% POFA
40% OPC + 60% POFA

Fig. 2: Scanning electron micrograph of POFA
Table 1: Physical properties and chemical composition of
typical palm oil fuel ash
Tests

POFA

Physical Properties
Fineness - Sp. surface area (m2/kg)
Soundness – LeChatelier method (mm)Specific Gravity

520
1
2.22

Chemical Analysis (%)
Silicon Dioxide (SiO2)
Aluminium Oxide (Al203)
Ferric Oxide (Fe2O3)
Calcium Oxide (CaO)
Magnesium Oxide (MgO)
Trioxide (SO3)
Sodium Oxide (Na2O)
Potassium Oxide (K2O)
Loss of Ignition (LOI)

43.60
11.40
4.70
8.40
4.80
2.80
0.39
3.50
18.00

28-day Strength Activity Index with OPC

112

Setting Time (min)
150
195
225
300
375
420
420

3.2 Workability of Concrete
Concretes with pozzolans usually have an improved
workability than those made with un-substituted Portland
cement at the same water content though there are exceptions.
There are numerous evidences that pulverised fuel ashes
improve workability of concrete because of their sphericalglassy particles. It has generally been thought that the
improvement of workability of concretes incorporating fly
ash is due to a ‘ball-bearing’ action of the particles. These finespherical particles are assumed to reduce the interference
between larger particles of the aggregate.
It is interesting to note that unlike fly ash concrete, POFA
concrete exhibits higher water demand. To attain the samelevel of workability, mixes containing 30% POFA required
higher water content than those containing only ordinary
Portland cement as the binder [16]. Table 3 illustrates that theslump of concrete for all mixes increased with the increase inwater-binder ratios. Although the slump of both OPC and
POFA concretes using superplasticiser was found to increase
reasonably, its response to POFA concrete at low
water-binder ratios was not significant. At higher
water-binder ratios, however, POFA concrete showed some
improvement but the slump values still remained below thatof concrete with OPC alone.
No single cause has so far been identified for the lower slump
value in POFA concrete. The cohesiveness of the mix due tothe higher surface area and high carbon content in the ash
could be responsible for the lower slump. Also, unlike typical
fly ash particles, the POFA particles seem to be
predominantly irregular although some particles with213

Table 3: Workability of OPC and POFA concreteType ofConcrete

Without
admixture

With 1%superplasticiser

4.1 Ash Quality and Development of Strength
The hydration of cement paste containing pozzolanicmaterials is complex in nature, and the factors which areresponsible for hardening of concrete with fly ash have, in
general, shown to influence the strength of concrete
containing palm oil fuel ash. Among others, type and amountof ash, curing of concrete are most critical. Like other fly
ashes the fineness of palm oil fuel ash has been shown to
influence the strength of concrete [14],[19]. Fig. 4
demonstrates that finer the ash, higher is the development ofstrength.

Slump of Concrete (mm) at
Water-Binder ratio of
0.4

0.45

0.5

0.55

0.6

OPC

0

0

20

30

60

POFA

0

0

10

10

30

OPC

40

50

80

120

180

POFA

10

30

65

100

165
28-Day Compressive Strength (MPa)Mix

4. STRENGTH AND DEFORMATION BEHAVIOUR

spherical shape appeared in the scanning electron micrographas illustrated in Fig. 2. Possibly because of such character,other pozzolanic materials like rice husk ash behave similarly[5],[17].
3.3 Bleeding Characteristics
Over the decades, extensive research has been carried out into
the causes and mechanism of bleeding in cement paste andconcrete. Among others, the properties of cementing
materials largely affect the bleeding characteristics, and, ingeneral, it is decreased by the increase in the fineness of thecement. The bleeding characteristics of concrete containing
various amount of palm oil fuel ash have been studied [16]and illustrated in Fig. 3. It can be seen that bleeding in POFA
concrete was much lower than in OPC concrete at all levels of
replacements. The accumulated bleeding water after 30
minutes, for example, in OPC concrete was 60ml; whereas
concrete with 10, 30 and 60% POFA bleeded only 35, 20 and
5ml respectively after the same period of test. Similar
observation goes with other pozzolanic materials like ricehusk ash and silica fume. The reason for the low bleeding
would possibly due to the fact that the extremely fine particles
attach themselves to the cement particles, reducing thechannels for bleeding, and leaving very little free water
available in the fresh concrete for further bleeding [18].

60
49.2

50
35.5

40
30

26.5

20
10
0
100% Passing 600 µm 100% Passing 300 µm 85% Passing 45 µm
Unground (Coarse) Unground (Medium)
Ground (Fine)

Type of Ash

Fig. 4: Effect of fineness of ash on strength of concrete
Although it has been possible to replace 30% OPC by POFA
without any loss of strength [16], replacement levels from 10
to 20% have been found to be optimum in producing highstrength concrete [19].
4.2 Effect of Curing on Strength Development
There has been a school of thought that the strength
development of concrete containing pozzolans at early ages isslower than that of concrete with only Portland cement. Likeother typical fly ash concrete POFA concrete has been shownto develop lower strength at early ages [16]; at later ages thecompressive strength of POFA concrete has been found to beslightly higher than the OPC concrete (Fig.5).

100% OPC
10% POFA
30% POFA
60% POFA

100
80

100
Compressive Strength (MPa)

Accumulated Bleed Water (ml)

120

60
40
20
0
0

20

40

60

80

100

60
40
OPC Concrete

20

POFA Concrete

0

120

0

Elapsed Time (min)Fig. 3: Bleeding characteristics of OPC and POFA concrete

80

100

200

300

400

Curing Period (Day)

Fig. 5: Strength development of concrete containing POFA214

28-Day Compressive Strength (MPa)Interestingly POFA concrete has been identified to be highly
sensitive to wet curing as the reactions of ash in concrete takea relatively longer time; prolonged weight curing appears to
be vital for the development of strength (Fig. 6). Insufficient
curing especially during early ages, therefore, seems to be
detrimental to the long term strength of POFA concrete.
Similar observations have been made by Abdullah et al. [20]on lightweight aerated concrete containing palm oil fuel ash.
70
60

POFA Concrete
52.15

52.9

53.5

OPC Concrete
55.25
50.2
45.7

50
40
30
20
10
0
Water

7-Day Water + Air

Air

Curing Regime

Fig. 6: Effect of curing regime on compressive strength of
OPC concrete and concrete with 30% POFA.
4.3 Influence of Admixture
Over the years there have been tremendous advances in thedevelopment of chemical and mineral admixtures. The use ofchemical admixture particularly superplasticisers of various
types in fly ash concrete has been emphasised since thebeginning of the wide application of ashes in structuralconcrete. Although fly ashes impart workability to a certain
extent, the use of superplasticiser has been shown to beindispensable for the improvement of strength and long-term
durability, especially when the amount of ash is large.
Short-term investigation on the compatibility of POFA
concrete with chemical and mineral admixture indicates(Table 4) that superplasticiser has a positive response on thedevelopment of strength of concrete.
Table 4: Strength development of concrete as influenced bychemical and mineral admixture
Concrete Mix

Control(100% OPC)
Double Blend(70% OPC+30%POFA)
Double Blend(70% OPC+30%POFA)
with 1% superplasticiser
Triple Blend(65% OPC+30%POFA+ 5% SilicaFume) with 1% superplasticiser

Compressive Strength
(MPa)
1
Day

7
Days

28
Days

17.55

37.40

46.60

14.70

35.10

45.50

16.85

36.25

50.60

16.00

37.15

49.70

Limited data on concrete with silica fume have also shown to
influence the strength of concrete by compensating the slower
strength gain at early ages although the overall performanceof silica fume in POFA concrete was not much satisfactory.
4.4 Modulus of Elasticity
Modulus of elasticity of concrete containing POFA has beenstudied by several researchers and has been described indifferent ways. In general, however, it has been shown to beclosely associated with the magnitude of the corresponding
compressive strength of concrete. Ishida [21] demonstrated
that modulus of elasticity of 20% POFA concrete was lower
at early ages, but slightly higher than that of control after one
year. Concrete containing 30% POFA exhibited similar trend
although the compressive strength was a bit lower than that of
the control specimen. The observation made is consistent
with the short-term research findings of Awal and Hussin [22].
In another experiment Sata et al. [19] found that the ground
POFA content up to 30% by weight of cementitious materials
had little effect on modulus of elasticity of concretes. With
the compressive strength value of about 90 MPa, it was foundthat ground POFA gave slightly lower modulus of elasticity
than those of control specimen. This lower value, as they
presumed, was due to the decrease in coarse aggregate
content of the concrete mixes. Tangchirapat andJaturapitakkul [23] experienced it in different ways. Although
the use of coarse and fine POFA did not influence the value,
the modulus of elasticity of ground POFA concrete increased
with the increase of compressive strength, and was found to
be about 7% higher than that predicted by ACI 318.
4.5 Shrinkage and CreepIn addition to the deformation due to load, a concrete
specimen exhibits other types of deformations – creep andshrinkage. Creep is the deformation when a concrete member
is subjected to a sustained load, and shrinkage is thecontraction suffered by concrete even in the absence of load.
Like that of other pozzolanic materials, the properties of
POFA have been shown to influence the shrinkage and creepof concrete. Along with modulus of elasticity, Ishida [21]carried out extensive study on shrinkage behaviour ofconcrete containing POFA over a period of one year. It has
been found that POFA concrete, in general, exhibited moreshrinkage value than the concrete with OPC alone.
Tangchirapat and Jaturapitakkul [23] also demonstrated that
concrete incorporating coarse ash did not have reduced
drying shrinkage; however, concrete containing fine palm oil
fuel ash exhibited 10-17% lower drying shrinkage than OPC
concrete.
Awal and Nguong [14] conducted short-term investigation on
shrinkage of concrete containing high volume palm oil fuelash. The measured values of shrinkage over a period of 28days, plotted in Fig. 7, reveal that the shrinkage strain of
POFA concrete was higher than that of concrete with OPC
alone. The magnitude of shrinkage of OPC concrete at 28-day,for instance, was 285 x 10-6. At the same time about 21%
higher value of shrinkage i.e. 347x10-6 was recorded for theconcrete with 50% palm oil fuel ash. The generalobservations made by Brooks and Neville [24] corroborates

215

40

600
R.H: 85 ± 5%; Temp: 27 ± 2oC
7 Days Initial Moist Curing

400
300
200
OPC Concrete
POFA Concrete

100

OPC Concrete
POFA Concrete

38
Temperature (oC)

Shrinkage (10-6)

500

36
34
32
30
0

0
0

5

10

15

20

25

20

30

40

60

80

100

120

Time After Casting (hr)Time (Day)the findings in this study. They have demonstrated that at aconstant water-cement ratio, a higher proportion of fly ash
leads to higher shrinkage magnitude by some 20%.
Unlike shrinkage, research findings on creep of concrete
containing palm oil fuel ash are not many. The observation
made by Ishida [21] suggests that the inclusion of POFA in
concrete did not influence much of creep properties. The
specific creep of concrete containing 30% POFA was foundto be 5% lower than that of control at 180 days. The creep
co-efficient of POFA concrete, as he observed, was about
11% lower as compared to that of concrete without ash.
5. ASPECTS OF DURABILITY
5.1 Resistance to Heat Rise
The hydration of cement compounds is exothermic i.e. heat is
generated within the concrete matrix during hydration. The
temperature rise of concrete due to hydration is largely
controlled by materials and mix properties, and byenvironmental factors. Like that of chemical admixture theuse of pozzolanic materials in reducing heat of hydration in
concrete is well established.
Experimental investigation conducted by Awal and Hussin[25] on resistance to heat rise demonstrates that the partial
replacement of OPC by POFA is advantageous particularly
for mass concrete where thermal cracking due to excessive
heat rise is of great concern. The development of temperature
due to heat liberation by the hydration process was measured
at the mid-depth of concrete blocks. Fig. 8 reveals that during
the initial period, the temperature rise for both OPC and
POFA concrete was almost equal. With time, a two-foldeffect of the partial replacement of OPC by POFA can bedetected. Firstly, concrete with POFA reduced the total
temperature rise and secondly, it delayed the time at which
the peak temperature occurred. Although the initial mixing
temperature of both the mixes was approximately the same,considerably more heat was evolved from OPC concreteduring the first day i.e. within twenty four hours after casting.
The peak temperature of 36.7oC, as obtained for OPC
concrete, was recorded at 20 hours. On the other hand, theFig. 8: Development of temperature in OPC and POFA concrete

highest temperature of 35.4oC was monitored in concrete with
30% POFA at 28 hours of casting.
5.2 Resistance to Chloride Penetration
Concrete, either plain or reinforced, can be harmfully affectedby chloride action. To protect from chlorides less permeable
concrete and adequate cover thickness are generally
considered. In addition to these precautions, however,utilisation of special cements and particularly incorporationof pozzolanic materials have been proved to be useful inminimizing the harmful effects.
Investigation on the resistance to chloride ion revealed (Fig.
9) that the depths of penetration of chloride ions in concrete
containing 30% POFA was much lower than in concrete with
OPC alone [26]. The pozzolanic behaviour of the ash causing
a decrease in permeability resulting from the refinement ofpore structure of the cement matrix has been attributed for thehigher resistance to penetration of chloride ion into concrete.
15
Depth of Penetration (mm)Fig. 7: Drying shrinkage of OPC concrete and high volume
POFA concrete

10

5
OPC Concrete
POFA Concrete

0
10

20

30

40

50

60

Immersion Period (hr)Fig. 9: Chloride penetration in OPC and POFA concrete
5.3 Resistance to Sulphate Attack
Among all chemicals, sulphates are the most significant inview of their attack on concrete. Sulphate attack on concrete
is a complex phenomenon that involves a series of primaryand secondary chemical reactions, some of which areinterrelated. The inclusion of pozzolanic materials inincreasing the resistance to sulphate attack is well recorded in216

Mortar Expansion (Microstrain)

the literature. Many pozzolans have been found that caneffectively double the service life of a concrete when exposed
to sulphate attack [27]. Although the composition of pozzolanthat influences the resistance of concrete to sulphate attack is
not fully understood, but in general terms it appears that
low-calcium fly ashes have been reported to be effective
mineral admixtures for combating sulphate attack on concrete[28]. The results on the expansion of mortar bars preparedfrom ordinary Portland cement and the one blended with
POFA have been presented in Fig. 10.
8000
0% POFA

6000

30% POFA

4000
2000
0
0

10

20

30

40

50

60

Exposure Period (Week)

Fig. 10: Expansion of mortar bars due to exposure in 10%
sodium sulphate solution
5.4 Resistance to Acid Attack
Cement based materials are subjected to acidic environment
in a variety of ways. The resistance of concrete to acid attack
can be increased either by artificial surface treatment [29] orby incorporating suitable pozzolans [18],[30]. Like concretewith other fly ashes of low-lime content, POFA concrete
exhibited a commendable resistance to acid attack [26].
Concrete specimens exposed to hydrochloric acid solution fora period of 1800 hours demonstrated (Fig. 11) that at all
periods of immersion the loss of weight of OPC concrete washigher than that of concrete with 30% POFA. The pozzolanic
behaviour as well as its low CaO content had beencharacterized for the resistance to such acidic environment.
30
25
Weight Loss (%)

5.5 Alkali-Silica Reactivity
The alkali-silica reaction, in its simple meaning, is an internal
chemical reaction between the hydroxyl ions derived from thealkalis and certain forms of reactive silica present in theaggregate of a concrete mass. The use of supplementary
cementing materials such as fly ash, slag and silica fume isone of the most popular solutions proposed to suppress the
expansion due to alkali-silica reaction [31],[32].
The extent of research in fly ash and alkali-aggregate reaction
has proliferated in the past decades. Despite higher alkali
content, laboratory tests following the South African
accelerated mortar bar test method demonstrated that POFA
has a good potential in suppressing expansion due toalkali-silica reaction [33]. Inevitably, the fineness of POFA
and its pozzolanic behaviour, identified responsible for its
unique resistance to other aggressive chemicals, acted infavour of its effectiveness in preventing expansion owing toalkali-silica reactivity.

20
15
10
OPC Concrete

5

POFA Concrete

0
0

300

600

900

1200

1500

1800

2100

Immersion Period (hr)Fig. 11: Comparative weight loss of OPC and POFA concrete
specimens immersed in 5% hydrochloric acid solution

5.6 Performance in Marine Environment
The use of concrete in marine structures goes back to theancient Romans and Greeks. Since that time numerous
marine structures have been built using concrete, and many of
the structures made with pozzolans are still in good condition.
However, the history of concrete in marine environments
demonstrates that no construction materials are perfectly
durable to sea water which presents one of the most corrosive
natural environments in the world. Most importantly concretein the marine conditions differs from the concrete in otherenvironments in that, under water, the material iscontinuously exposed to the action of complex sea water, andabove water, to spray or airborne salt and other mechanicalload [34]-[36]. With the increasing marine activities, the useof pozzolanic materials such as fly ash, silica fume, andnatural pozzolans in concrete contribute to the formation of adenser binder which inhibits the migration of the sea water
into concrete. Indeed, due to the higher pozzolanicity, theperformance of these materials in marine environment havebeen well marked [37]-[40]. The durability performance
study of concrete containing palm oil fuel ash in marine
environment was carried out [41], [42] by installing theconcrete specimens in sea under the jetty of Marine Police
Station along Johor Straits near the Malaysia-Singapore‘causeway’. Table 5 illustrates the performance behaviour ofconcrete at various exposure periods.
Results summarised in Table 5 reveal that all the concrete
specimens had an overall very satisfactory appearance at allages of inspection. The strength data obtained here in, on thewhole, suggest that the performance of concrete containing
palm oil fuel ash in sea is quite satisfactory. Penetration data,
given in the table also reveal that at all periods of exposure thedepth of penetration of chloride ions into OPC concrete washigher than in concrete with POFA. No significant
differences in carbonation, however, have been noticed in
OPC and POFA concrete specimens. At early exposure
periods, the depth of carbonation was almost zero. At laterages, only a small amount of carbonation in both OPC and
POFA concretes was detected.

217

Table 5: Characterization of ageing performances of concrete
exposed to marine environment

Tests

VisualObservation

Compressive
Strength(MPa)
FlexuralStrength(MPa)
Chloride
Penetration(mm)

Carbonation(mm)

Types of Concrete

Exposure
PeriodOPC

POFA

2 years

Almost intact
with few bore
holes on thesurface of theconcrete
specimen

Same asOPC
concrete

6 months

56.8

58.1

1 year59.9

59.5

2 years

60.7

61.9

6 months

6.85

7.70

1 year6.95

7.00

2 years

6.90

7.50

6 months

21.0

13.5

1 year

23.5

17.0

2 years

28.0

20.5

6 months

0.0

0.5

1 year1.5

1.5

2 months

2.0

1.5

6. CONCLUDING REMARKS
While the whole world is going green, the constructionindustry in both the developed and developing world is also
going green. Indeed, the sustainable development has become
the core of green technology i.e. transition or transformationof all activities towards greener future through green actions.
Due to the increased interest in green building concepts andpractices, more supplementary cementing materials are
expected to be seen in the market. The experimental findings
and the observations made so far, suggest that the inherent
pozzolanic properties of palm oil fuel ash have resulted insignificant quality improvements including enhancement ofstrength and durability. These high performance
characteristics have positioned POFA as a suitable material
when resistance to aggressive chemical environments andlong-term strength characteristics are called for. It remains,
however, for the construction industry to adopt and advancethe material technology to its full utilization towards thegreen concrete construction.
7. ACKNOWLEDGEMENTS
The author wishes to acknowledge the technical support
received from the Department of Structure and Materials,
Faculty of Civil Engineering, Universiti Teknologi Malaysia.
The ‘research university grant’ acquired from the Research
Management Centre (RMC) of the university (Code:
Q.J130000.7122.01J81) is also gratefully acknowledged.

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[33] Awal ASMA and Hussin MW, “The effectiveness ofpalm oil fuel ash in preventing expansion due toalkali-silica reaction”, Cement and Concrete Composites,
vol. 19, no. 4,1997, pp. 367-372.
[34] Browne RD and Baker AF, “The performance ofstructural concrete in a marine environment”,
Developments in Concrete Technology-1, Applied
Science Publishers Ltd., London, 1979, pp. 111-149.
[35] Marshall AL, Marine Concrete, Blackie and Son Ltd.,
London, 1990.
[36] Al-Rabiah AR, “Concrete durability consideration for
the King Fahad Causeway – a case study”, In Proc. of the
4th International Conference on Structural Failure,
Durability and Retrofitting, Singapore, 1993, pp.
199-205.
[37] Hoff GC, “Concrete for offshore structures”, Advancesin Concrete Technology, 2nd ed., CANMET, Ottawa,
1994, pp. 83-123.
[38] Mehta PK, “Concrete in the Marine Environment,
Elsevier Science Publishers Ltd, 1991.
[39] Malhotra VM, Carette GG and Bremner TW,“Performance of high-volume fly ash concrete at Treat
Island, Maine, In Proc. of the 6th International
Conference on Durability of Building Materials andComponents, Omiya, Japan, 1993, pp. 1011-1020.
[40] Roper H, Sirivivatnanon V and Baweja D, “Long-term
performance of Portland and blended cement concretes
under marine conditions”, In Proc. of the 3rd International
Conference on Durability of concrete, Nice, France, ACI
Publication SP-145, 1994, pp. 331-351.
[41] Hussin MW and Awal ASMA, “Influence of palm oil
fuel ash on strength and durability of concrete”, In Proc.
of the 7th International Conference on the Durability of
Building Materials and Components, Stockholm,
Sweden, 1996, pp. 291-298.
[42] Awal ASMA and Hussin MW, “Concrete in marine
environment: influence of palm oil fuel ash on strengthand durability”, In Proc. of the Civil and Environmental
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Bangkok, Thailand, 1999, pp. 11149-11156.

219

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Behavior of Collapsible Loessic Soil After Interparticle Cementation
Arrua P., Aiassa G., Eberhardt M., Alercia Biga I., Grupo de Investigación y transferencia en Geotecnia,
Estructuras y Fundaciones (GIGEF), Facultad Regional Córdoba, Universidad Tecnológica Nacional. Argentina

ABSTRACT: The water content, shape and roughness of soil
particles are related to the strength and stiffness of loessic soil
samples in remolded state. When soil sample are in undisturbed
stage, the cementation level between particles by the presence ofwater-soluble carbonates, govern the macroscopic behavior. Soil
stiffness depends on the rigidity of the link. To control themechanical collapsible behavior usually compaction andstabilization is made with cementations agents. This work presents
the hydration process of cement without mixing water and how
geotechnical index and parameter are modified. The variables that
influence the resistance have been evaluated by unconfinedcompression tests. Results showed that resistance depend with theinitial water content of hydration. Optimal conditions for generation
cementing bridges are established.

Keywords: micrograph, hydration, strength, unconfinedcompressive
1. INTRODUCTION
In the construction of geotechnical structures, theimprovement of soil properties is very common in Argentine
1-4. Many studies in the literature are related to the use of
cement in the improvement of engineering properties ofsoils5-8 . The compacted soil and soil mixtures must be
constructed with local soil, so soil behavior is relevant todesign. The Córdoba city is located in the geographical centerof Argentina. Loess soils are usually found in arid or semiarid
climates and their physical and chemical characteristics
depend on their geological origin 9. In nature the loess has
an open structure, low unit weight, highly dependent on
external conditions. The purpose of the addition of cementing
agents is to stabilize the silty loess due to lack of technicalqualities for application in construction works.
Environmental conditions such as temperature and humidity
significantly influence the strength characteristics ofsoil-cement mixtures 10. The amount of cements
incorporated, is generally related to the strength and stiffness
assigned to the material at the stage of design, however thisdefinition may be inadequate if the environmental conditions
during the stage of hydration are not considered in the final
strength of material 11, 12. This paper presents thechemical composition, principal geotechnical properties,
particles size and relation between different kinds of loess inArgentine. An electron microscope has been used to establish
the chemical composition of local cement. Different samples
have been built to study the behavior of loess-cement
mixtures. To quantify the stress strain relation, unconfinedcompression test have been made. Time of curing,temperature of curing, initial water and cement contents, arethe variables that govern behavior. The results allow

establishing the roughness and sphericity of the particles. It
also identifies the link between them. The results showed thatstress-strain curves modify its initial elasticity modulus withvariations of cement content.
2. MATERIALS
2.1 Loessic soil
Loess is non stratified aeolian deposit, and probably mostabundant Quaternary deposit on land 13. It consists of siltwith some small fraction of clay, sand and carbonate. The
deposits are up 30 m in thick en the Missouri and Rhine Rives
Valleys, more than 180 m thick in Tajikistan, 330 m thick innorthern China 14 and 2-80 m thick in Argentine. The mostimportant deposit in Latin America is located in central areaof Argentine.
These kinds of soils have been deposited during the Upper
Pleistocene and Holocene and cover much of the province of
Córdoba. Reference 15 has been described how, lightyellow or brown in color, some times with a reddish or grey
tinge. Usually, most researchers agree that the material
originates from the Andes Mountains, with volcanic
mineralogy agents. Recent contributions on the knowledge ofthe origin and evolution of the loess are presented in 9 and16. Locally, the soil has got mineral content from Small
Hills of Córdoba Province. Grain-size distribution includes
fine sand (1%-10%), silt (50%-80%) and clay (2%-15%) thathave been deposited by wind action in areas of low energy.
Fig. 1 presents typical grain size distribution. The chemicalcompositions of loess are mainly SiO2, Al2O3, Fe2O2 y CaO.
Chemical compositions of the soil from different regions ofArgentina are shown in Table I. Note that close similarity
exists between those loess, except for the percentages of CaO(which reflect the influence of locally degradation of Cordoba
Small Hills). However the chemical composition isremarkably uniform, suggesting a common regional source.
In natural conditions the water content (%) ranges from 12.7
to 23.0, the dry unit weight (kN/m3) is 12.5 to 13.5, Specificgravity is 2.66 to 2.67, Atterberg limits are: liquid limit 23% –30% and plasticity index 4.2% – 4.9%. Dynamic Cone
Penetration Index [mm/blow] is 18–24, for Mohr-Coulomb
failure criterion the friction angle in triaxial test (CD) is 27º –30º, and cohesion [kN/m2] in triaxial test (UU) is 15 – 40.
Poisson modulus rages from 0.3 to 0.35, and elasticity
modulus from unconfined compression test (MN/m2) is 1.5–8.
From oedometer test, the secant modulus for 100kPa in
MN/m2 is 2.0 – 7.0. Usually yield pressure in oedometer test
(a: vertical direction; b: horizontal direction) is (a) 60kPa –221

MgO: 2.79%, K2O: 1.97%, Na2O: 0.59%, SO3: 5.10%, TiO3:
0.47%. The most important difference with soil is content of
SiO2 and CaO.

280kPa, (b) 100kPa – 180kPa.
US standard sieve

Hydrometer

Percent finer by weight (%)

100

80

60

40
Depth: 1.0m
Depth: 2.0 m
Depth: 3.0 m

20

0
10

1

0.1

0.01

0.001

Grain size (mm)Gravel Coarse

Sand
MediumSilt or Clay

Fine

Fig. 1. Gradation result for loess soil
2.2 Cement
The chemical composition of Normal Portland Cement CP40
and particle morphology have been determined by a LEO
1450 VP Energy Dispersive Spectroscopy analysis (EDS) and
Scanning Electron Microscope (SEM). The results shows that
ordinary cement manufactured in Malagueño-Córdoba have
SiO2: 15.42%, Al2O3: 4.60%, Fe2O2: 3.35%, CaO: 65,71%,

2.3 Soil cement mixtures
The mixtures were designed to evaluate the performance inUnconfined Compression (UC). A review of literature about
the testing of laboratory samples of soil-cement indicates thattest results are highly dependent on sample preparation. This
work includes variations in cement content, time of curing,temperature of curing, initial water content and compactionlevel.
The cement contents (C.c.) used, expressed as a percentage ofthe dry soil weight, were: 2.5%, 5.0%, 7.5%, 10% and 15%.
The soil was dried in an oven at 105 ºC for 24 hours, wassieved and recovering the material passing sieve #40. Water
was added in the proportions established at each case as
mixing water. The samples were generated by compaction in
the mold at constant dry unit weights (approx. 14 kN/m3). The
manufacture of the samples was performed in three equal
layers scarification between them. Were classified andarranged in different curing conditions. Variabletemperatures were taken from -20°C to 120°C.

Table I: Oxides soil components
SiO2 Al2O3 TiO3 Fe2O2
58.35 16.64--6.6967.22 14.35--6.8159.20 13.60--4.3059.86 17.40--4.8062.70 15.00--6.0057.16 17.28--5.4366.01 16.22
0.88
5.30

CaO
8.98
5.16
6.50
3.08
2.80
2.83
2.85

MgO
2.75
2.13
1.10
1.17
1.90
1.67
1.62

K2O
2.82
2.93
2.20
1.70
1.88
3.68
1.90

Na2O
2.11
1.41
3.10
1.97
1.40
8.35
1.97

59.00

17.00

---

5.87

3.05

2.55

1.56

1.38

---

14.00

0.80

4.60

0.09

1.50

3.00

1.90

3 SHAPE AND COMPOSITION OF PARTICLE
The shape and connection between particles are an inherent
soil characteristic that play a mayor role in macroscopic
mechanical properties. The morphology describes at large
scale the level platy or sphericity, and texture in smallerscales reflect the local roughness.
In this work, the chart from 22 is used to compare individual
grains for visual estimation of roundness and sphericity. The
chart from 23 is used for shape characterization. The
particle morphology, roughness and roundness have been
determined by a LEO 1450 VP Scanning Electron
Microscope (SEM) and the components of loess soil with an
Energy Dispersive Spectroscopy analysis (EDS).
The microscopic morphology of loess was shown in Fig. 2. It
could be seen from these SEM results that the surface of loess

H2O------6.04
4.32--3.14

Reference
Soil used in this work. Cordoba City
Córdoba City 17Provincia de Córdoba 18Baradero, La Plata 13Miramar. Buenos Aires 13
Buenos Aires 15
La Pampa 19Valles Preandinos subtropicales; Provincia de
5.80
Tucumán 20--- Llanura Chaco Oriental 21material are subrounded and rounded with 0.7 sphericity
coefficient and a 0.5 to 0.7 roundness coefficient. The size
particles distribution is homogenous. The silt particles are thebiggest and the smaller size are clay particles. Fig. 2 shows
the zone to zoom and analyzed in Fig. 3. The cement presence
affects the structure inherent of soil, because small particles
stick to larger particles in stable connections. SEM photos ofloess, cement, soil cement mixture at C.c: 2.5% and 15% arealso shown in Figs. 3. Fig. 3(a) shows that particle size taken
in consideration for comparison with mixtures, has size inupper limit of silty soil (< 75m and < 45m). It can beseen that particle has angular and subangular shape with somesmall silty particles. Also appears some clays particles. The
Fig. 3(b) presents Portland cement particles. It shows that
predominant grain sizes are smaller than silt and with size
between 1m to 10 m. It could be seen from these SEM

222

that sphericity coefficient is 0.3 to 0.5 and roundness
coefficient is 0.3-0.5. From SEM photo, it is found that theparticles of cement are smaller than loessic soil.

to establish links with others silt.

Fig. 4. SEM photo of mixture C.c.: 2.5%
Fig. 2. SEM morphology of loess

There are cementation between smaller particles and big
silt particles, but we do not find cementation between big silty
particles. The low cement content may be responsible for thelinks absence.
Fig.5 shows the SEM for a cement content of 15%. Whenthe cement content are high (15 % of C.c.) were able toestablish stable cementation and links between different silt
particles. This phenomenon may be responsible for increased
stiffness and strength of mixtures samples. The ellipses areas
indicate links between silt and hydrations products.

(a)Fig. 5. SEM photo of mixture C.c.: 15%(b)Fig. 3. SEM photos. (a) Natural loess soil. (b) Normal
Portland cement
As shown in Fig. 4 mixture after 21 cementation days withcement content of 2.5% do not shows relevant difference with
Fig. 3(a). When the cement content is low, stick on the
cement particles some of silty soil, forming clusters that failAnalysis of Energy Dispersive Spectroscopy is presented
in a schema three dimensional (Fig. 6). It shows the minerallevel present in cement, loessic soil and a mixture of soil
cement with C.c: 15%.
The figure shows that the content of Si decreases withincreasing of cement content (C.c), while calcium (Ca)decreases with decreasing of cement content. The content of
Ca are 35.50%, 12.91% and 5.34% for cement, mixturesoil-cement (15% of Cc), and loess respectively. The content
of Si are 5.46%, 18.63% and 22.78% for cement, mixturesoil-cement (15% of Cc), and loess respectively.

223

22.78
18.63

35.50

5.46

12.91

5.34

V)
Energy (kFig. 6. Energy Dispersive Spectroscopy test (EDS). Variation of Si and Ca

4 MECHANICAL TESTFor unconfined compression tests (UC) a mechanical press
was used, instrumented with a load cell with a capacity of
50kN and a digital comparator for recording displacementswith a precision of 0.001 mm, to a constant deformation rateof 2.4 mm/min. UC tests is used to evaluate the stress–strain
characteristics and the stiffness properties of theSoil-Cement-Mixture (SCM).
Fig. 7 shows the typical failure of specimens with C.c: 15%
and natural loess soil (C.c: 0%). The addition of cement,
causes an increase stress and strength. SCM samples have atypical inclined failure plane (f) for defining cohesion andfriction angle in Mohr-Coulomb failure criterion. In loesssamples without cement addition (C.c: 0%) there is anincrease of sectional side due to material ductility. The failure
stress is considered as maximum stress reached before
softening in stress-strain curve. It does not take into accountthe level of deformation reached.
The results obtained from 5 UC tests with unit weight: 14
kN/m3, time of curing: 14 days, temperature of curing: 20 ºC,
and initial water content: 21 % is shown in Fig. 8.
The tendency on stress-strain curves show that the initialmodulus increases with increasing of cement content. The
tendencies are not linear and deformation reaches 2.5%.
Cement content of 2.5% have a unconfined compressive
strength of 270 kPa, but a cement content of 15% can increase
resistance to 9628 kPa. Cementation, cohesion and suctionlevel, modify cohesive parameter from Mohr-Coulomb
failure criterion. (Fig.2 (b)), it grows to 1564 kPa for C.c:
15%, 1463 kPa for C.c.: 10%, 987 kPa for C.c: 7.5%, 208 kPa
for C.c: 5.0%, 71.95 kPa for C.c: 2.5%. For each C.c specifiedin paragraph 2.3, has been modified water content. This
allows establishing its influence in the development ofresistance. The specimens were built to 14 kN/m3 dry unit
weight.


ilur
Fa

f

f

el

ine

f
f
f

2f

f



(a)
(b)
(c)
Fig.7: Typical failure. (a) Soil-cement (C.c.: 15%) (b)Theoretical behavior model (c) Natural loess soil
Fig. 9 shows the maximum compressive strength of
concrete mixtures and the initial water content. Samples builtwith Cc: 2.5%, presents a maximum resistance to wi = 19%,
for Cc: 5.0% at wi = 23%, Cc: 7.5% at wi = 19%, Cc: 10.0%
at wi = 21 %, and Cc: wi = 15% to 23%. The results show thatthe peaks of resistance are obtained for water contents
between 19% and 23%, average 21%. Outside this interval,
the unconfined compressive strength decreases significantly.
This behavior can be associated with deficiency or excess inavailable water in the soil matrix during the hydration
process.

224

10.0

5

Axial stress (MPa)8.0Compressive strength c(MPa)

Cement content:
15.0%
10.0%
7.5%

6.0

5.0%
2.5%
Failure

4.0
2.0
0
0

0.5

1.0

1.5
2.0
Axial strain (%)

2.5

Deferred gainstrength

4

80

3
2

Instant gain hydration
resistance
Hydration water content: 21%
Cement content: 9%

60

40

1

20

0

0
0.01

3.0

100

0.1

1

10

100

Time of curing tc (days)

Fig.8: Stress-strain curves unconfined compression test

Fig. 10: Compressive strength and time or curing
This phenomenon prevents the hardening material
generation and the cement links between particles. The lowest
resistance obtained for tests, is presented for low
temperatures. Possibly because of increased volume of water
during freezing, and decrease in water available forhydration.

Maximum

20
Minimum
15.0%

7.5%

5.0%

10

10.0%

15
Cement content

3Compressive strength c (MPa)

25

2.5%

Initial water content wi (%)

30

5
0
0

2

4

6

8

10

12

Compressive strength c (MPa)Fig. 9: Maximum compressive strength for different initialwater content
A cement content of 9% has been defined to study thecuring time influence and curing temperature influence. Fig.
10 presents the relationship between unconfined compressive
strength and curing time, with 21% initial water content and
20 °C curing temperature.
At early ages less than 15 days, developed the mostimportant percentage of resistance, which reaches 3.5 MPa.
In semilogarithmic scale, the relationship between curing
time and the resistance is linear.
After 15 days, there is a strength improvement less than 20%of the maximum strength.
Fig. 11 presents results that relate to the curing temperatureand unconfined compressive strength. The samples werestored during the curing period in different environmental
conditions. The samples were cured in an oven, refrigerator
and laboratory environment. Temperature was controlled.
For temperatures below 0 °C, the maximum strength
reached is 2.0 MPa. The optimum is between 20 ºC and 30 ºC.
The results showed that unconfined compressive strength
decrease (40%), for temperature extreme conditions (100 ºC).
Decreased resistance to low and high temperatures may
occur by inhibition of water available for hydration duringfreezing effect or vaporization effect respectively.

2
Maximum
strength1

0-20

Initial hydration water content: 21%
Cement content: 9%
Age: 15 days

0

20
40
60
80
Curing temperature Tc (ºC)

100

120

Fig. 11: Compressive strength vs. temperature of curing

5 CONCLUSION
This paper has presented a study on loess material andnormal Portland cement and revised the importance ofinterparticle bridges and their influence on the mechanicalperformance of the material. Water content, unit weight,
cements content, time of curing and temperatures of curing
have been studied. The principal results are as follows: The unconfined compression shows a maximum
independent of cement content when initial watercontent range is 19% to 21% (hydration water).
 The temperature range that triggers the hydration of
cement and improves the unconfined compressive
strength ranges from 20 ºC to 30 °C.
 Extreme temperatures (high or low) will reduce strengthof samples in 60%.
 80% of unconfined compression strength is achieved225during the first 15 days of hydration.
 Mixtures with Cc: 7.5% or higher have appropriate
physical and mechanical properties for use in various
civil infrastructures. Highlights the possibility ofmaking bricks for building houses.

6 ACKNOWLEDGMENT
This work was partially funded by the National University
Technological, Regional Córdoba Faculty (Argentina), by the
National Agency of Scientific and Technological Promotionof Argentina and Ministry of Science and Technology(Government of Córdoba). The authors thank the staff of
GIGEF.

[17] Arrúa, P.; Aiassa, G. and Eberhardt, M. 2011. Loessic soil stabilizedwith cement for civil engineering applications. International Journal of
Earth Sciences and Engineering. ISSN: 0974-5904 (to be published)[18] Ricci J. 1966. El loess de Río Tercero y el probable origen de losmallines (Córdoba). Terceras Jornadas Geológicas Argentinas, Bahía
Blanca.
[19] Arens, P. 1969. Algunos paisaje geoquimicos de la región pamepeana,
Reunión Argentina de la Ciencia del Suelo, Santa Fe, Argentina, Actas
V.
[20] Camino, M., 1988. Estratigrafía y evolución paleoambiental durante elCuaternario del Valle de La Sala. Unpublished Thesis, Facultad de C.
Naturales, Universidad de Tucumán, Tucumán, Argentina.
[21] Morras, H., 1996. Composición y evolución de la fracción limo grueso
de suelos del Chaco meridional argentino. Congreso Geológico
Argentino, Actas XIII Vol. IV: 263-265.
[22] Krumbein W., and Sloss L. 1963. Stratigraphy and Sedimentation, 2nded. W.H. Freeman, San Francisco.
[23] Powers M. 1953. A new roundness scale for the sedimentary particles.
Journal of Sedimentary Petrology, Vol. 23, pp.117-119.

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Arrúa, P; Aiassa, G. Eberhardt, M. and Alercia, I. 2011. Estabilizaciónde suelos loessicos mediante la incorporación de un aglomerante
mineral. 14th Pan-American Conference on Soil Mechanics andGeotechnical Engineering. Toronto, Ontario, Canadá 2 - 6 October
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[2] Eberhardt, M.; Arrúa, P.; Aiassa, G. y Terzariol, R. 2011. Optimizaciónde las obras de cierre para el control de inundaciones en el noreste
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226

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

A Study on Hardened Soil Brick Using Liquid Type and Powder Type
Hardening Agent
Jin, Sung Ki. Jung, Seung Yong. Kim, Tae Ho., Overseas Business Team, EGE(Expert Group for Earth &Environment) Co. Ltd.

ABSTRACT: The study is to produce hardened soil bricks, blocks,
boundary stones, or other objects. It uses hardening agent (liquid- or
powder-type) made with eco-friendly ingredients. Contrarily to the
traditional bricks, baked in the furnace, it is possible to makehardened soil bricks and blocks from site soil, whose particles are
unevenly distributed, without applying heat to them. Therefore, apurpose of this study is to provide a method of the hardened soil
brick production which is able to reduce costs and meet thedemanding quality standard by utilizing the site soil. To achieve this,laboratory mixing test and uniaxial compression test have been
performed regarding specimen that mixes liquid and powder type
hardening agent with the site soil not meeting the standard ofgrain-size distribution. Input percentage of chemical components forthe agents is also evaluated by XRF. Through the tests, improvement
effect on the hardening agents may be verified, and moderate mixing
ratio of them may be drawn.

Keywords: Hardened Soil Brick, Powder Type Hardening
Agent, Liquid Type Hardening Agent, OMC(Optimum
Moisture Content), Maximum Dry Density(γdmax)
1. INTRODUCTION
Lately, the demands for economical and satisfactory methodsof road base treatment and brick production have been
increasing in developing countries such as China, Southeast
Asian countries, and African countries. The fundamental
reason for that is the short supply of quality sand and crushedstone in some regions or the financial difficulty to supply
them.
Currently, the road bases in developing countries are treatedwith concrete, but a technology to utilize the site soil isneeded to reduce the expenses for the disposal of surplus soil
and the treatment itself. On the other hand, bricks areproduced by the conventional method, of which the site clay
is collected with human hand then put in a mechanical former
to shape clay bricks, and then the clay bricks are dried under
the sun and burned in the furnace. Bricks produced in thisway vary in quality since the heat a brick receives in the
furnace differs depending on the spot it is placed. Also, themethod caused deforestation when firewood is used tooperate the furnace and resulted air pollution when fossil fuelis used.
Other than burned clay brick, cement brick is out in themarket. The approximate cement - sand - crushed stone ratio
for cement brick production is 1:2:4. There are two ways todecide the quantity of water added, which have a great impact
on the solidity of brick. One follows the approximate

cement-to-water ratio of 1:10. The other is to go by the call of
the site engineer. It is done by grabbing the mixture of
aggregate, cement, and water with his fist and judges the
O.M.C. (Optimum Moisture Content) by the sensation in his
hand, which gives rise to varied quality. Also, if cement isused by itself, heavy metal eruption is possible and thestrength is increased limitedly unless crushed stones areadded. To prevent these problems, the additives are mixed
with the cement. There are two different kinds of typicalhardening agent, liquid type and powder type. The liquid typeis suitable for sandy soil, and the powder type clayey soil. For
these agents, input percentage of chemical components isevaluated by XRF(X-ray fluorescence) analysis in order tosurvey the harmful components.
This paper presented two different ways for the production of
the hardened soil bricks, which is elastic, economical, andeco-friendly and exceeds by far the target strength, 8 N/mm2(81.6kg/cm2), with high durability and no heavy metal
eruption. The first of the two ways mentioned is to mix
cement and sandy soil and add a little quantity of liquid type
hardening agent. The second is to mix clayey soil and cement
and add power-type hardening agent.
A purpose of this study is to provide a method of the hardened
soil brick production and the road base treatment, which ischaracterized by the aggregate of the site soil, cement, and thehardening agent mixed by the agitator, vibrated, pressed byoil pressure, and molded. Another purpose is to present thematerials and mixing ratio of liquid and powder type
hardening agents for the production of soil brick and the roadtreatment using the site soil (sandy soil, clayey soil) which isunevenly distributed.
2. FIELD OF STUDY AND METHOD
2.1 Experimental Method
Considering the unfavorable condition for the site soil quality
in some countries, soil sampling for this study was conductedat Jebudo beach in Incheon-si, Korea, the silt and sand sample
of which does not meet the standard of grain size distribution,
Coefficient of Uniformity (Cu), of 4 to 6 [4].
The quality control standard was established according to theresult of the tests done to different types of hardening agentsfor road base treatment and brick production. Strength test
and pre-mixture test was performed identically to each type,
and then the agents which satisfy the criteria from those tests
were tested in detail. Two different types of hardening agent,in broad, were developed and the improvement to each soil

227

type accordingly was shown after a series of tests. The first is
liquid type which is mainly composed of chemicals, and theother powder type which is mainly mineral materials[2]-[3]-[7].
2.2 Used Hardening Agents
2.2.1 Liquid type hardening agent
Liquid type hardening agent used for this study is inorganic,and Sodium silicate (Na2SiO3), Magnesium chloride (MgCl2),
Potassium chloride (KCl), Calcium chloride (CaCl2), andSodium carbonate (Na2CO3) are the additives for thehardening agent. The chemical reaction formulas among theadditives are as follows:
CaCl2+ Na2SiO3→CaSiO3+2NaCl(1)

KCl + Na2SiO3→KSiO3+2NaCl(2)

MgCl2+ Na2SiO3→MgSiO3+2NaCl(3)

H2O+ Na2SiO3→SiO2H2O+2NaCl(4)physical properties of silica fume is that its particles are very
minute. Above 95% of its particles is less than 1 micrometerin size. When silica fume is added to cement slurry, the fineparticles of silica fume fill the space between cement particles
and this phenomenon is called Micro Filling. Because Micro
Filling substitutes the pores bigger than 1 micrometer in size
to innumerable microscopic pores, using silica fume increases
the coherent force among soil particles and the strength ofsolidified body. Also, silica fume, which has a highpercentage of amorphous Silicon Dioxide (SiO2), turns into
highly reactive pozzolan material in soil. It chemically reacts
to Calcium Hydroxide (Ca(OH)2), the product of chemical
reaction between Portland cement and water(H2O), and forms
the binder called Calcium Silicate Hydrate(C-S-H). Thisbinder especially improves the strength of solidified body. Its
high amorphous Silicon Dioxide (SiO2) content andminuscule particles cause silica fume to be highly reactive
especially in the early phase. Furthermore it is known thatpowerful dispersing agent or chemical admixture, known as
the water reducing agent, increases effectiveness of silica
fume [5]. The chemical equation for powder type hardening
agent of this study shows Pozzolanic reaction and as follows.
Particularly silica fume, SiO2, is artificially supplied topromote formation of the Calcium Silicate Hydrate(C-S-H)
binder [5]-[6].

The by-products of this process, which are Calcium silicate(CaSiO3), Potassium chlorate (KSiO3), Magnesium
silicate(MgSiO3), and Silicon dioxide Hydrate(SiO2H2O), areassumed to complexly react to the main component of normal
cement such as SiO2, Al2O3, CaO, and Fe2O3, with the chief
ingredient of site soil like SiO2, Al2O3, Fe2O3, CaO to cause
solidification [1].
The Input Percentage of the Component of Liquid Type
Hardening Agent is as Table Ⅰ.

C3S, C2S(in cement) +H2O→Ca(OH)2+H2O↓
Add SiO2 to the product↓
The formation of C-S-H(Calcium Silicate Hydrate)

Table Ⅰ. The Input Percentage of the Component of Liquid
Type Hardening AgentThe Input Percentage of the Component of Powder type
hardening agent is as TableⅢ.

Description

SiO2

CaO

MgO

Na2O

K2O

Percentage(%)

50.08

3.02

3.08

2.25

4.08

Description

SO3

Cl

CHO

OthersPercentage(%)

1.68

18.61

16.50

0.70

Table Ⅱ. The Standard Mixing Ratio for Liquid Type Soil
Brick

Sandy Soil

Liquid Input(kg/m3)
1.5

(6)
(7)

Table Ⅲ. The Input Percentage of the Component of Powder
type hardening agent

The Standard Mixing Ratio for Liquid Type Soil Brick which
satisfies the target strength of 8N/mm2 (81.6 kg/cm2) for
hardened soil brick in this study is specified in Table Ⅱ
below.

Description(5)

Cement Input(kg/m3)400

Description
Percentage(%)
Description
Percentage(%)

SiO2

Al2O3

Fe2O3

CaO

90.09

1.05

1.43

1.33

MgO

1.77

Na2O

K 2O

SO3

Cl

Others
0.26

0.50
2.76
0.68 0.13
The Standard Mixing Ratio for Powder Type Soil Brick
which satisfies the target strength of 8N/mm2 (81.6kg/cm2)
for hardened soil brick in this study is specified in Table Ⅳ
below.
Table Ⅳ. The Standard Mixing Ratio for Powder Type Soil
Brick

2.2.2 Powder type hardening agent
Silica Fume which is used for powder type hardening agent inthis study is micro-fine powder that is the by-product of
producing silicon and ferrosilicon alloys in electric arc
furnace. Although it is essential in the production ofhigh-strength concrete, the use for it as an additive to thehardening agent for brick is very insignificant. One of the228

Description

Cement(kg/m3)

Clayey Soil

520

Silica Fume
(kg/m3)26 (5% of
Cement Weight)

Lignosulfonate(kg/m3)
1.3 (0.25% of
Cement Weight)

3 RESULT
3.1 Result of Soil Property Test
Table Ⅴ. Result of Soil Property Test
Description
Depth (m)Moisture Content (%)
Density, ρs (g/cm3)
Liquid Limit (%)
Plastic Index
D
Compaction
USCS

γdmax

(t/m3)
OMC
(%)

Sand
0.30
0.30
1.4
1.1
2.692
2.694
NP
NP
-

0.30
29.7
2.697
27.9
2.1

0.30
29.4
2.697
29.1
2.7

1.62

1.60

-

-

Silt

18.0

-

17.0

-

SP

SP

ML

ML

Maximum Dry Density (γdmax)
Optimum Moisture Content (OMC)

1.62
18.0

t/m3
%

Fig. 3 Result of Compaction Test (Sand)

The average Uniform Coefficient (Cu) of the site soil samples
for this study was recorded 2.89 for sandy soil and 2.56 forclayey soil, which means the soil particles are unevenly
distributed. According to the compaction test, sandy soil
sample showed Maximum Dry Density (γdmax) of 1.62 t/m3,
and Optimum Moisture Content (O.M.C) of 18.0%. Also,
clayey soil sample showed Maximum Dry Density (γdmax)
of 1.60 t/m3, and Optimum Moisture Content (O.M.C) of
17.0%.
The result of grain-size distribution test for sand and silt isshown in Fig. 1 and 2. The result of compaction test for sandand silt is also shown in Fig. 3 and 4.
Maximum Dry Density (γdmax)
Optimum Moisture Content (OMC)

1.60
17.0

t/m3
%

Fig. 4 Result of Compaction Test (Silt)
3.2 Result of the Uniaxial Compression Strength Test

Fig. 1 Grain-Size Distribution Curve (Sand)

Fig. 5 Cement Input vs. Uniaxial Compression Strength of
Liquid Type Soil Brick

Fig. 2 Grain-Size Distribution Curve (Silt)

229

such as SiO2, Al2O3, CaO, and Fe2O3 and cause thesolidification.

1. Silica Fume and lignosulfonate are used for powder type
hardening agent as the main additives.
Also, silica fume turns into highly reactive pozzolan material
in soil and forms the binder called Calcium Silicate
Hydrate(C-S-H). When it is mixed with cement and acts bothFig. 6 Cement Input vs. Uniaxial Compression Strength of
Powder Type Soil Bricks

as water reducing agent and binder for strength increase.

According to the uniaxial compression strength test carried

2. Sodium silicate and various additives are used for liquid

out to soil brick specimen, liquid type hardening agent is

type hardening agent.

more effective to sandy soil sample, and powder type to silt

These materials act with the normal cement and cause thesoil sample. As for liquid type sand brick, the maximum

strength increase.

strength was yielded when the input of the agent was
2.5kg/m3, and as for powder type clay brick when the input of
the mineral-consisting powder type agent reaches 8% of thetotal cement weight. Also, the mixture is to be compacted to-2% of Optimum Moisture Content(O.M.C.) in the case of
liquid type sand brick, and +2% of O.M.C. in the case of
powder type silt brick. The optimum mixing ratio for liquid
type sand brick is 2.5kg/m3 of liquid hardening agent and3. For sandy soil, the standard mixing ratio for liquid type soil
brick which satisfies the target strength of 8N/mm2 (81.6kg/cm2) for hardened soil brick in this study is specified asfollows; Liquid Input: 1.5(kg/m3), Cement Input: 400(kg/m3)

4. The input percentage of the component of liquid type

400kg/m3 of cement. The optimum mixing ratio for powder

hardening agent is as follows; SiO2: 50.08(%), CaO: 3.02(%),

type sand brick is when the input of powder type mineral

MgO: 3.08(%), Na2O: 2.25(%), K2O: 4.08(%), SO3: 1.68(%),

additive is 8% of the total cement weight, and the cement

Cl: 18.61(%), CHO: 16.50(%), Others: 0.70(%)

input is 520kg/m3, and this satisfies the strength standard for

The input percentages of the component of powder type

cement brick as shown in Fig. 5 and 6 above [8].

hardening agent are as follows; SiO2: 90.09(%), Al2O3:
1.05(%), Fe2O3: 1.43(%), CaO: 1.33(%), MgO: 1.77(%),

4 CONCLUSION
Liquid type hardening agent used for this study is inorganic,and Sodium silicate (Na2SiO3), Magnesium chloride (MgCl2),
Potassium chloride (KCl), Calcium chloride (CaCl2), andSodium carbonate (Na2CO3) are used as the additives for thehardening agent.
The by-products of this process, which are Calcium silicate(CaSiO3),

Potassiumchlorate(KSiO3),Magnesium

silicate(MgSiO3), and Silicon dioxide Hydrate(SiO2H2O),

Na2O: 0.50(%), K2O: 2.76(%), SO3: 0.68(%), Cl: 0.13(%),
Others: 0.26(%)

5. For clayey soil, the standard mixing ratio for powder type
hardening agent which satisfies the target strength of 8N/mm2(81.6kg/cm2) is specified as follows; Cement: 520(kg/m3),
SilicaFume:

26(kg/m3)(5%of

CementWeight).

Lignosulfonate : 1.3(kg/m3)(0.25% of Cement Weight)

react complexly with the main components of normal cement230

5 REFERENCES
[1]
[2]
[3]
[4]
[5]
[6]
[7]
[8]
[9]

Moon Han Kim, “Study on Development of Soil Cement Brick andvalidity for its application”, Journal of the Architectural Institute of
Korea, Volume 24, No. 94, 1980, pp 25-32.
Korea Institute of Industry and Technology Information, “Study on
optimum treatment of by-product gypsum”, 1998.
Ministry of Construction and Transportation, “Study on the reuse of thecasting foundry fly ash - Final Report of Research and Development
Project for 1996”, 1998.
Sang Kyu Kim, “Soil Mechanics - Theories and Applications”, Cheong
Moon Gak Publishers, 1992.
Korea Concrete Institute, “Concrete Structures”, 1995.
Silica Fume Association, “Sillica Fume User's Manual”, U.S.
FedDeperal Highway Administration, 2005.
Byung Sik Chun, “A Study on the Surface Soil Stabilization on Marine
Clay by the Hardening Agent”, Journal of Ocean Engineering andTechnology, Volume 15, No. 1, 2001, pp 92-97.
Korean Standards Association, “Korean Industrial Standards Concrete Brick(KS F 4004)”, 2003.
“The Deep Mixing Method(Principle, Design, Construction)”, Coastal
Development Institute of Technology, Japan, 2002, pp 37-50.

231

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Effect of Salt Content in Soil on the Sorption Capacity of Heavy Metals
Lerd Kerdchaiyaphum and Pongsakorn Punrattanasin
Research Center for Environmental and Hazardous Substance Management and Department of Civil Engineering
Faculty of Engineering, Khon Kaen University, Thailand
Email: lerdkerd@gmail.com, ppunrattanasin@gmail.com
Abstract
In the Northeastern part of Thailand, saline soil can be easilyfound in abundance in all areas. Salts can be directly
extracted from the topmost layer of soils. This study aimed
to determine the sorption values of contaminations on soil
with a variety of salt content. A soil sample in this study wascollected from Khon Kaen University. The soil is known as
loess or silty fine sand (SM). The amount of contaminations
adsorbed in soil was measured using the batch equilibrium
test. Chemical analyses were performed with an Atomic
Absorption Spectrophotometer (AAS). The equilibrium
conditions required are around 48 hours of shaking between
loess with heavy metals. The results also show theadsorption isotherm and the maximum adsorption capacityof soil with different contents of salts.
Keywords: Contaminations, Saline soil, Sorption value,
Isotherm, Loess
1. Introduction
Due to technologies in the Northeastern part of Thailand
growth during many decades, industrial has been increased.
The results in a discharge of waste into environmental overthe limit. The environmental problems are severe anddifficult to resolve. One result is leakage of heavy metals incontaminated water and soil [2]. Heavy metals are some ofthe deposit in the soils. Because soils are heterogeneous,numerous studies have focused on the interaction of several
heavy metals with different soil constituents [4]. The
Northeastern region of Thailand covers an area of
168,855.50 km2. The top most soil layer found in all areas inloess layer or wind-deposited soil (Fig.1). The thickness ofloess ranges from a few meters to more than six meters. In
the Northeastern part, saline soil can be also easily found inabundance in all areas, especially in the top and middle
parts. Thus, this research aims to investigate the effect of saltcontent in soil on the adsorption capacity of heavy metals by
the loess.
2. Material description
The material used in this study is loess which is collected
and excavated from Khon Kaen University. The soil can beclassified as silty find sand (SM) in the Unified Soil
Classification System (USCS). The basic, index andengineering properties of the loess are tabulated in Table 1.

Figure 1. Loess deposit in the Northeastern part of Thailand.
Table 1. Loess properties.
Properties
Soil classification (USCS)
Specific gravity
Natural dry unit weight (kN/m3)
Natural water content (%)
Liquid limit (%)
Plastic limit (%)
Plasticity Index (%)
Optimum moisture content (%)
Maximum dry unit weight (kN/m3)
Coefficient of Permeability (cm/s)

Loess
SM
2.60
15.30
5-12
16.0
13.0
3.0
10.0
21.1
2.80 x 10 -6

Heavy metals chosen in this study was only copper (Cu),zinc (Zn) and nickel (Ni) which was obtained by dissolving
nitrate. Tables 2, 3 and 4 summarized properties of copper
nitrate, zinc nitrate and nickel nitrate, respectively.
Selections of Cu, Zn and Ni as heavy metals were based on
its low toxicity and relative ease involved in handling,
disposal and also all are easily found in all areas. All three
are the common heavy metals which are discharged fromvarious and common industries [1].
Table 2. Properties of copper nitrate.
CompoundCopper NitrateCu(NO3)2
FormulaMolecular weight (g/mol)
295.65
Density (g/cm3)
2.07
Solubility (g/100 ml)
137.8

233

Table 3. Properties of zinc nitrate.
CompoundFormulaMolecular weight (g/mol)
Density (g/cm3)
Solubility (g/100 ml)

Zinc NitrateZn(NO3)2
297.492.065
137.8

Table 4. Properties of nickel nitrate.
CompoundNickel Nitrate
FormulaNi(NO3)2
Molecular weight (g/mol)
290.79Density (g/cm3)
2.05
Solubility (g/100 ml)
94.2
Figure 3. Shaking soil and solution.
3. Experimental methodology
In this study batch adsorption test is designed as a tool tostudy the equilibrium adsorption of soil suspensions withcontaminants [5]. In batch adsorption test, chemical solution
and disaggregated state are brought into contact [3]. The
main objective of this test is to find the adsorption capacityof heavy metals by loess and to find the effect of salt content
in soil on the sorption capacity of heavy metals.
3.1
Batch adsorption test
The equilibrium contact time was firstly determined andinvestigated by batch adsorption test. Soil sample was loess(SM). This sample was washed to clear as much as
contaminant away first after that, sized it by taking throughby the sieve No. 200. Heavy metals used in the experimentwere Copper Nitrate, Zinc Nitrate and Nickel Nitrate withthe concentration of 100, 200, 300, 400 and 500 mg/L. A
constant mass of oven soil powder of 2.5g and 50 ml of
metal solution were added into a set of 120 ml plastic bottles(Fig 2.). The soil solution ratios were then shaken at 150
rpm for 0.5, 1, 3, 6, 12, 24, 48 and 72 hours by using ahorizontal shaking machine (Fig 3.). After particular shaking
time the solutions were filtrated though 0.45μm of filter toseparate the soil and liquid solution. Liquid solution wasthen analyzed the chemical concentration by using Atomic
Absorption Spectrophotometer (AAS) (Fig 4.).

Figure 4. Atomic Absorption Spectrophotometer (AAS).
3.2
Salt content on the adsorption test
The salt was mixed with loess with different contents. The
salt contents were starting from 10% up to 90% by weight ofsoil. Heavy metals used in the experiment were the same asin the previous section. A constant mass of oven soil powderof 2.5g, percentage of salt by weight of soil and 50 mg/L ofmetal solution were added into a set of 120 ml plastic
bottles. The soil solution ratios were then shaken at 150 rpm
for 48 hours by using a horizontal shaking machine. After
that, the experiment was the same as batch adsorption test.
4. Experimental result
4.1
Bath adsorption test
Batch adsorption test were carried out with the aim ofobserving the equilibrium contact time of copper, zinc andnickel by loess. The results are shown in Figs. 5-7 forcontact time of copper, zinc and nickel by loess,respectively. The equilibrium sorption of copper, zinc andnickel were found in the 48 hours. After achieving
equilibrium condition if soil solution was kept in still state, itwould not cause any sorption or leaching.

Figure 2. Soil and solution.

234

600

C = 100 mg/L
C = 300 mg/L
C = 500 mg/L

500

3

C = 200 mg/L
C = 400 mg/L

2.5
Csolid (mg/g)

C (mg/L)

400
300
200

1.5
1

100

0.5

0

0
0

12

24

36
48
Time (hr)

60

72

600

C = 100 mg/L
C = 300 mg/L
C = 500 mg/L

500

Cu

0

84

Ni

50 100 150 200 250 300 350 400 450 500 550
Ceq (mg/L)

After that, the results of chemical analyzed were plotted
between the adsorbed concentration of soil ( Csolid ) versus

C = 200 mg/L
C = 400 mg/L

the equilibrium concentration of liquid ( Ceq ) in order to see

400

the highest tendency that soil can adsorption. For loess withall three solutions, the curves are in Fig. 8. All results show
the non linear relationship between Csolid and Ceq thus the

300
200
100
0
0

12

24

36
48
Time (hr)

60

72

84

non linear models were implied in the analysis. The result
will be plotted at the graph for find the slope and theintercept form to compare with Langmuir (Fig. 9). Langmuir
sorption isotherm has been developed on the concept that asolid surface possesses a finite number of sorption sites andmathematically expressed as

Csolid 
Figure 6. Sorption capacity of zinc by loess.
600

C = 100 mg/L
C = 300 mg/L
C = 500 mg/L

500

 Ceq1   Ceq(1)

where α = the adsorption constant related to the binding
energy (L/mg) and β = the maximum amount of contaminantthat can be absorbed by the solid (mg/g). To determine thevalues of α and β, a relationship of Ceq versus Ceq / Csolid is

C = 200 mg/L
C = 400 mg/L

400
C (mg/L)

Zn

Figure 8. The sorption isotherm of zinc, copper and nickel
by loess.

Figure 5. Sorption capacity of copper by loess.

C (mg/L)2

plotted. The determined values of α and β based on thecalculation from Figs. 9 and 10 for different isotherms areshown in Table 5. It was found that the copper has thehighest sorption capacity by considering the α and β values
from its isotherm.

300
200
100
0
0

12

24

36
48
Time (hr)

60

72

84

Table 5. Langmuir isotherm parameterαβSoil Solution(L/mg) (mg/g)
Copper 0.0085
3.132
Loess
Zinc
0.00781.976
Nickel
0.0067
2.543

Figure 7. Sorption capacity of nickel by loess.

235

Correlation
Coefficient (R2)
0.997
0.999
0.995

350

300

Cu

Zn

Ni

250
200
150

150
100

50

50

0

C = 200 mg/L
C= 400 mg/L

200

100

0

C = 100 mg/L
C= 300 mg/L
C= 500 mg/L

250

Ceq (mg/L)

Ceq/Csolid (g/L)

300

050 100 150 200 250 300 350 400 450 500 550
Ceq (mg/L)

0% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

Percentage of salt

Figure 11. Sorption capacity of copper by loess mixing
with salt.

Figure 9. Langmuir isotherms of toxic metals by loess.

400
350

C = 100 mg/L

C = 200 mg/L

C= 400 mg/L

C= 500 mg/L

C= 300 mg/L

Ceq (mg/L)

300
250
200
150
100
50
0
0% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

Percentage of salt

Figure 12. Sorption capacity of zinc by loess mixing
with salt.

Figure 10. Langmuir isotherm’s parameters.
4.2
Salt content on the adsorption capacity
The results of chemical analysis of soil with heavy metals
with different contents of salt were plotted between theequilibrium concentration of liquid ( Ceq ) versus theloess mixing with salt ranked from the highest were copper,
nickel and zinc, respectively.

350

C = 100 mg/L

C = 200 mg/L

C= 400 mg/L

C= 500 mg/L

C= 300 mg/L

300

Ceq (mg/L)

percentage of salt in order to see the highest tendency thatsoil can adsorption. The results are shown in Figs.11-13.
The sorption capacities of toxic metals by loess mixing withsalt were found that the percentage of salt staring from 60%
by weight can provide the equilibrium concentration ofliquid ( Ceq ). The sorption capacities of heavy metals in

400

250
200
150
100
50
0
0% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

Percentage of salt

Figure 13. Sorption capacity of nickel by loess mixing
with salt.

236

Then, the results of chemical analyzed of loess mixed salt
were plotted between the adsorbed concentration of soil
( Csolid ) versus the percentage of salt. The salt was mixed
with loess. It was found that a salt content starting from 60%
by weight can provide the highest sorption capacity. The
results are shown in Figs.14-16. The sorption capacities ofheavy metals in loess ranked from the highest were copper,
nickel and zinc, respectively.
8
7

8
7

C solid (mg/L)

C = 200 mg/L

C= 400 mg/L

C= 500 mg/L

C= 300 mg/L

C solid (mg/L)

6
5
4
3
2

C = 100 mg/L
C= 400 mg/L

C = 200 mg/L
C= 500 mg/L

C= 300 mg/L

1060% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

5Percentage of salt

4

Figure 16. The relationship between Csolid and percentage ofsalt from the test of nickel.

32
1
00% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

Percentage of salt

Figure 14. The relationship between Csolid and percentage ofsalt from the test of copper.
7
6

C = 100 mg/L

C = 200 mg/L

C= 400 mg/L

C= 500 mg/L

C= 300 mg/L

5

C solid (mg/L)

C = 100 mg/L

4
3
2
1
0
0% 10% 20% 30% 40% 50% 60% 70% 80% 90% 100%

Percentage of salt

Figure 15. The relationship between Csolid and percentage ofsalt from the test of zinc.

5. Conclusions
This research presents sorption values of contamination on
loess which can be found in the top most layers in Khon
Kean University. A possible method of reducing the amountof toxic in the river is also proposed by adding and mixing
the salt content into the original soil. Batch adsorption test
can give an estimate for maximum possible amount ofcontaminant that can adsorbed by soil. The following
conclusions are results in this study from batch adsorption
test;1) The equilibrium sorption of copper, zinc and nickel
were found in the 48 hours.
2) The sorption capacities of heavy metals in loessranked from the highest were copper, nickel andzinc, respectively. The highest sorption capacity byconsidering the  and  values from Langmuir
isotherm.
3) The salt was then mixed with the loess from Khon
Kaen University in order to investigate the effectof salt content starting from 60% by weight can
provide the highest sorption capacity. The sorption
capacities of heavy metals in loess ranked from thehighest were copper, nickel and zinc, respectively.
6. Acknowledgment
This research was funded by Research Center for
Environmental and Hazardous Substance Management,
Khon Kaen University. The authors wish to acknowledge
and grateful thank to Asst. Prof. Dr. Arthit Neramittagapong
and Miss Wimonporn Iamamornphonth of the Department ofChemical Engineering, Khon Kaen University.

237

7.

References[1] Alloway, B.J., 1995. “Heavy metal in soil”, 2nd ed. Blackie
academic and professional.
[2] Bedient, P.B., Rifai, H.S. and Newell, C.J. 1994.
“Groundwater Contamination”, Prentice-Hall.
[3] Fetter,
C.W.
1999.
“ContaminantHydrogeology”,
Prentice-Hall.
[4] Selim, H.M. and Sparks, D.L., 2001. “Heavy metals release insoils”, Liwis Publishers.
[5] Tanchuling, M.A. 2005. A study on sorption and dispersion ofheavy metals in clay using column leaching test. Ph.D. Thesis,
Tokyo Institute of Technology, Tokyo, Japan.

238

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Manufacture of Housing Construction Materials with Jute Polymer
Composite (Jutin)
Mubarak A. Khan*, Fahmida Parvin, Jahid M M Islam, Md. Zakaria Hossain1
Institute of Radiation and Polymer Technology
Bangladesh Atomic Energy Commission, P. O. Box 3787, Dhaka 1000, Bangladesh
1
Department of Environmental Science and Technology, Mie University, Japan
Conventional heavy andbrittle building materials
such as stones, bricks, mortar, granite etc. do notabsorb shock waves but they amplify them, causing
more destruction. The material of construction to
be used in earthquake prone areas should be suchthat they absorb & reduce seismic energy [1].
Jute has high cellulose content and low micro-fibril
angle which are desirable properties of a fiber to be
used as reinforcement in polymer matrices. Jute isone of the most common natural fibers having hightensile modulus and low elongation at break. If the
low density (1.45 g/cm3) of this fiber is taken into
consideration, then its specific stiffness andstrength are comparable to respective quantities ofglass fiber [2, 3]. In spite of these above-mentioned
advantages, jute fiber – like other natural fibers –exerts some difficulties while used asreinforcement in non-polar polymer matrices.
Being polar and hydrophilic in nature, jute fiber
exhibits poor interfacial adhesion with hydrophobicpolymer matrices. To overcome these kinds ofbottlenecks, many attempts, such as physical andchemical treatments, lead to changes in the surface
structure and surface energy of the fibers.Such aneffort was made to prepare a composite withimproved mechanical properties using radiationinduced jute-urethane polymer system. The resin
matrix was prepared under gamma radiation using
urethane acrylate in the presence of Nvinylpyrrolidone, ethyl hexyl acrylate andtrymethylol propane triacrylate. Some additives
such as acetic acid, acrylamide, urea, talc, andtitanium oxide were incorporated into theformulation [4]. In this context, hessian cloth (juteproduct) was coated with urethane pre-polymer
with different formulations in the presence ofplasticizers under UV radiation. Tensile propertiesof the composites were found to increase. It is also
indicative that simulated weathering and soil
degradation tests show the biodegradable nature of
the prepared composites [5]. A good correlation
was found between composite impact damping and

ABSTRACTWith jute fiber (hessian cloth) reinforced unsaturated
polyester (USP) resin along with additives; composites
were prepared by simple hand lay-up technique at room
temperature (25°C). The prepared composite has been
given shape of corrugated tin, so termed as “JUTIN”(jute + tin). The mechanical properties (such as tensile
and bending properties) of the composites were measured
with the extent of jute contents (20-40%wt) and foundhighest at 30% fiber content. To improve the fiber-matrix
adhesion, jute fiber is subjected to chemical treatment
with curing and coupling agent, and physical treatment
with UV radiation of different intensities. Composite
made of chemically modified jute fabrics shows bettermechanical performances than that of the untreated or
UV treated composites. The two treatments together
provide synergistic effect on the properties of thecomposite. The treated composites also show the best of
charpy impact strength. Storage and loss modulus as wellas cantilever tests of the composites were
performed.About35%weight of the composite is forjute.The tensile strength of the composite is 147MPa,
bending strength is 150MPa and impact strength is 68
kJm-2and will be stable more than 50 years.
Keyword: Jute, Composite, Natural Fiber, Jutin,
Constriction materials

1. INTRODUCTION
Among destructive and terrifying natural calamities,
earthquakes measuring intensity morethan 5.0 on
the Richter scale can cause immense damage to thebuildings and structures, dams, bridges, railways,
communication systems etc [1]. Earthquakes arecommon in Japan, one of the world's mostseismically active areas. The country accounts forabout 20 percent of the world's earthquakes ofmagnitude 6 or greater. Recently a temblor struck
the region of the Pacific where a magnitude-9.0
quake hit on March 11, triggering a huge tsunami.
At least 23,000 people were killed or left missingin those disasters, which destroyed hundreds ofhomes, offices and factories in northeasternJapan.As the occurrence of earthquakes isunpredictable, there is an ardent need to design andconstruct the buildings and structures to reduce thedamages, thus saving precious human lives.

239

yarn toughness for the jute-epoxy composites [6].
Two monomers such as 2-hydroxy ethyl methyl
acrylate (HEMA) and 2-ethyl hexyl acrylate (EHA)
were successfully used as novel coupling agents forjute fabric (hessian cloth) polypropylene composite.
The mechanical properties of the resulting
composites increased as a result of surface
treatment of the jute fabrics [7]. Polyester resin wasused as matrix material for composite fabrication.
The mechanical properties (tensile and bending
strengths) of the surface modified jute fabrics
reinforced
polyester
composites
improvedsignificantly [8]. The unsaturated polyester resin isquite useful for industrial and civilian world. As a
low-cost, rigid, high strength-to-weight material,
one can find its products in the form of mechanicalparts, pipes, tanks, electronic gears, etc.They canbe cured to give insoluble, infusible solid plastics
through a free radical curing process. Organic
peroxides are employed as free radical initiators
while tertiary aromatic amines and some organic
metal salts, such as cobalt naphthenate, are used as
curing promoters if needed Tertiary aromatic amineand cobalt naphthenate can significantly reduce thedecomposition temperature of peroxides via
chemical reduction processes [9]. The objective of
this research work is to prepare low cost lightweight, durable, earthquake and cyclone tolerance
housing materials and so on likewise products withsustainable local technology.

apower strength of 2 kW. The speed of themanicure was 4 meter/minute for each pass of thesubstrate under the lamp by maintaining different
UV radiation intensities expressed by number ofpasses.
2.3.Composite Fabrication
Composites were fabricated using a simple handlay-up technique. The working surfaces weretreated with releasing waxes to facilitate easy
removal of samples from the mold surfaces. Cobalt
napthenate(catalyst) and MEKP (curing agent)
were mixed thoroughly with USP at various
formulations before each operation. At the
beginning of fabrication, a gel coat with 2% MEKP
was uniformly brushed into the finished side of themale and female parts of the mold. After 1h, whencuring of gel coat was completed, each layer of thefiber was pre-impregnated with formulations madeof USP. The impregnated jute samples were thenplaced one over another as a sandwich. This
sandwich was placed into a mould. Both parts ofthe mold were tightened by screw-bolt and allowed
3 h for total curing (composite fabrication). The
composites were cut into rectangular pieces of
equal size (120×100×3mm3) for different tests. All
results are taken as the average of five samples foreach testing.
2.4.Mechanical Tests
The tensile and bending strength of the composites
were measured according to DIN 53455 and DIN
53452 standard methods by a universal testingmachine (Hounsfield S Testing Series, UK) with an
initial clamp separation of 20 mm and a cross-head
speed of 10 mm/min. Charpy impact strength of thecomposite was determined by an impact tester(MT-3016) according to the DIN EN ISO 179
standard in the flat wise, unnotched mode. The test
samples were conditioned at 25°C and 50%
relativehumidity for several days before testing andall the tests were performed under the same
conditions.The result of each test is taken as theaverage value of five samples

2. MATERIALS AND METHODS
2.1. Materials
Jute fabric was used as the reinforcing agent. The
matrix polymer was a commercial unsaturated
polyester resin (EPOLAC G-153ALX) containing
1.5% cobalt napthenate solution, 6% cobalt catalyst
as promoter, and 36% styrene as diluent. Methyl
ethyl ketone peroxide (MEKP) and wax were used
as curing agent and mold releasing agent
respectively.
REOLOSIL
fumed
silica(aerosealpowder)was used as filler. A substituted
aromatic tertiary amine was used as coupling agent.
2.2. Methods
Surface Modification
Hessian cloth was cut into rectangles (12 ×10 cm2)and temporarily fixed in a long square sized plate(50 ×50 cm2) where UV radiation could be given
together to six equal sized rectangular samples.
Then the samples were subjected to UV radiation(254–133 nm) using an irradiator (UV minicureMe-200, ISTTechnik, Germany), which delivers

3. RESULTS AND DISCUSSION
The effect of jute content (%wt) on the mechanicalproperties of the resulting composites is studied
here and shown in Figures 1 and 2. It was foundshows poor mechanical properties due to poor fiber
population and low load transfer capacity to one
another. levels of fiber content (such as 20%), thecomposite As a result, stress gets accumulated
atcertain points of the composites and highly240

TS and BS (MPa)localized strains occur in the matrix. Atintermediate levels of loading (30%), thepopulation of the fibers is just right for maximum
orientation and the fibers actively participate instress transfer. So, the mechanical properties of thecomposite reaches maximum. At high levels(suchas 40%) of jute content, the non homogeneous
fiber matrix adhesion becomes prominent whichleads to agglomeration among the fibers and stresstransfer gets blocked [10]. As a result, themechanical properties of the composite again
decreased. The composite of the optimized jutecontent (30%w/w) showed 205% increase in tensile
strength (TS), 141% in tensile modulus (TM),
226% in bending strength (BS) and 195% in
bending modulus (BM) than that of the resin based
composite.

45
40
35
30
25
20
15
10

65
60
55
TS and BS (MPa)

50

40
35
30
25

TS

BS

20
0

100

200

300

400

Number of UV Passes

Fig.3 Effect of UV radiation on tensile and bending
strength of jute-based composites

Figure 2. Effect of jute content on tensile and bending
modulus of composites

TS

20

25

30

BS

35

40

Jute content (%)

Fig.1 Effect of jute content on tensile and bending strengthof composite

1.6

TM and BM (GPa)

45

1.5
1.4
1.3
1.2
TM

1.1

BM

1
20

25

30
35
40
Jute Content (%)
Fig. 2. Effect of jute content on tensile and bending
modulus of composites

241

Fig.4. Effect of UV radiation on tensile andbending modulus of composites
The effect of UV radiation on the mechanicalproperties of jute based composites wasinvestigated and presented in Figures 3 and 4. Jutecontent in the UV treated composite is maintained
at about 30%. The UV intensities of different
passes (200-400 pass) were exposed to jutesurfaces.1 UV pass indicates 1 metre of UV
exposure in the machine. It was found that UV
intensity of 200 pass showed the highest
mechanical properties. Mechanical properties, suchas tensile and bending properties increase up to 200
pass, which gives 68% increase in TS, 78% in TM,
58% in BS, and 57% in BM relative to untreatedjute-based composites. The TS, BS, TM and BM of
the control composite were found 29.45 MPa,
38.35 MPa, 1.35 GPa and 2.54 GParespectively.
The increase of mechanical properties of thecomposite with increasing UV radiation may be
due to the intercross-linking between theneighboring cellulose molecules, which results inthe strength of jute fabrics. It is observed from
Figures 3 and 4 that mechanical properties of thecomposite increase with UV pretreatment up to acertain limit and then decrease due to the two
opposing phenomena, namely, photo cross-linking

and photo degradation that take place
simultaneously under UV radiation. At lower doses,
free radicals are stabilized by a combination
reaction and, as a result, photo cross-linking occurs.
The higher the number of active sites generated on
the polymeric substrate, the greater the grafting
efficiency. But at higher radiation, the main chain
may be broken-down and polymer may degrade
into fragments and, as a result, mechanicalproperties were found to decrease after certain UV
doses. An intense radiation results in a loss ofstrengths and a reduced degree of polymerization isobserved [10].
Jute fabrics were further treated with a coupling
agent (0.1 to 2.0 %) at the stage of preimpregnating the fibers with USP resin to show its
effect on to the composites. Figs 5 and 6 show theeffect of coupling agent on the properties of thecomposites.

TM and BM values than that of the control
composite.
Table 1 shows the synergistic effect of the twotreatments – coupling effect and UV treatment – on
the properties of the composite. It was found that
TS, BS, TM and BM values of UV+CA treatedcomposite were 96 MPa, 117 MPa, 4.2 GPa and 5.5
GPa respectively. The values are about 226, 205,211and 116% higher than that of the control
composite. The free radical and cross linking
mechanisms of the two effects (UV +CA) may bemainly responsible for the increase of mechanicalproperties of the composite.
Table 1.The effect of UV+CA on the properties ofcomposite
SampleUV+CA
treatedcomposite

70

TSand BS(MPa)

60

TS
(MPa)
96 ±
2.2

Properties
BS
TM
(MPa)
(GPa)
117±
4.2 ±
2.5
0.11

BM
(GPa)
5.5 ±
0.15

50
40

The charpy impact strength (IS) of the control andtreated composites were studied here. The IS
values of UV, CA and UV+CA treated and control
composites were found 35, 39 and 44 and 20 kJ/m2
respectively. The effect of glass fiber (GF) in themechanical properties of UV + CA treated jutefabrics composites was investigated and is shownin Figures 7and 8.
The weight fractions of glass fiber content in the
jute/GF hybrid composites varied from 5 to 100%.
The incorporation of glass fiber into jute fabrics up
to 75% helps the hybrid composites show bettermechanical properties over the other treatedcomposites. The values of TS, BS, TM, BM and IS
of the hybrid composites at 75% GF content were
found 140 MPa, 150 MPa, 6.55 GPa, 7.12 GPaand
68 kJ/m2respectively. The incorporation of GF
increases the reinforcement of jute fibers into thematrix. In a hybrid composite, the mechanicalproperties of composites are mainly dependent on
the modulus and percentage elongation at break of
individual reinforcing fibers. The increase inmechanical properties through the addition of glass
fiber to jute can be explained on the basis of higher
modulus and elongation at break of the glass fiber[11], whereas the extensibility of glass is low
compared to the jute fiber.

30
BS

20

TS

10
0

0.5

1

1.5

2

Concentration of CA (%w/v)

Fig. 5. Effect of coupling agent on tensile andbending strength of composites.

Fig. 6. Effect of coupling agent on tensile andbending modulus of composites
The coupling agent with a concentration of 1.0 %performed the best of mechanical properties. It isalso indicative that coupling agent treatedcomposite showed much better mechanicalproperties than that of the control composite. It was
found that 1.0% coupling agent treated composite
showed 90, 74.7, 77.77 and 57.48% higher TS, BS,242

TS and BS (MPa)

160
150
140
130
120
110
100
90
80

TS

0

20

40

BS

60

80

100

120

Glass fiber (% w/w)

Fig.10. The effect of different treatments on
the modulus of composites

Fig. 7. The effect of glass fiber content on tensile
and bending strength of jute/GF hybrid composite

5. Aging properties of Composite (JUTIN)
5.1. Thermal aging
Table 2.The effect of thermal aging on tensile strength of
JUTIN
Sample
Jutin

160
TS

BS

IS

120
Properties

100
80
60
40
20
0

Un

UV

CA

CA+UV

Tensile strength (MPa) 15 days
OºC
4ºC
50ºC
70ºC
116±2.0 99.8±1.9 93±1.82 78±1.8

The JUTIN was subjected to undergo different
temperature treatments at 15 days time periods toinvestigate the effect of thermal aging on tensile
strength of JUTIN. It was found that the strength ofcomposite increases with the decrease of
temperature. The TS decreased to about 19.0 and
3.0 % at aging temperature of 70ºC and 50ºC
respectively. Whereas, the incat 4ºC and 0ºC, TS
increased about 4.0 and 21.0% respectively.
5.2. Water aging
There is very negligible amount of water uptake(<1%) within 3 months. There is no change inmechanical properties within 3 months.
5.3. Normal weathering aging
The mechanical properties increased up to 7-10%
within 6 months and remain unchanged up to 12
months.
5.4. Acceleratedweathering aging
To study the weathering effect on the mechanicalproperties of Jutin, the samples of jutin were
exposed under simulated weathering tester from Qpanel Co. (model QUV, USA). The weathering
testing was performed in alternating cycles ofsunshine over 4h (65º±2ºC) and condensation for
2h (45º±2ºC). This aging test was carried out for
600 h.Owing to this test the losses of mechanicalproperties more or less zero up to 300 hours andabout 10% up to 600 h.

Fig. 8. The effect of glass fiber content on tensile
and bending modulus of jute/GF hybrid composite
4. Comparative Studies
Figures 9 and 10 shows the comparative data ofdifferent treatments on the properties of jute andjute/GF based composites. It was found that theglass fiber on the UV + CA treated jute based
composite performed the best mechanicalproperties.

140

Room
temp.
96

CA+UV+GF

Composites
Fig. 9. The effect of different treatments on
the strength of composites243[5] Khan M.A., Uddin M.K. and Ali K.M.I. (1997).
Degradable jute plastic composite.Polymer
Degradation and Stability; 55:1-7.
[6] Gassan J. and Bledzki A. K. (1999).
Possibilities for improving the mechanicalproperties of jute/epoxy composites by alkali
treatment of fibers.Composites Science andTechnology; 59:1303-1309.
[7] Khan M.A., Hinrichsen G. and Drzal L.T.
(2001). Influence of novel coupling agents onmechanical properties of jute reinforced
polypropylene composites. Journal of Materials
Science Letters; 20:1711-1713.
[8] Mohanty A.K, Khan M.A. and Hinrichsen G.
(2000). Influence of chemical surface modificationon the properties of biodegradable jute fabricspolyester amide composites. Composites:Part A;
31:143-150.
[9] Khan M.A., Ganster J. and Fink H. P. (2004).
Natural and man-made cellulose fiber reinforced
hybrid polypropylene composites.5th Global Woodand Natural Fiber Composites symposium, inKassel, Germany, April 27-28.
[10] KuangW. and Richardson, A. (2007).A New
Amine Promoter for Low-temperature Cure of
MEKP Initiated Unsaturated Polyester Resin
Systems. Composites Research Journal; 1(4): 9-13.
[11]Abdullah-Al-Kafi, Abedin M. Z, Beg M. D. H,
Pickering K. L and Khan M. A ( 2006).Study on
the Mechanical Properties of Jute/Glass Fiberreinforced UnsaturatedRadiation Polyester Hybrid
Composites: Effect of Surface Modification by
Ultraviolet Radiation. Journal of Reinforced
Plastics and Composites; 25: 575-588.
[12] Khan, M. A. and Hasan, M. M. (2004).
Surface Modification of Natural Fibers by Photografting and Photo-curing, In: Mital, K. L. (ed.),Polymer Surface Modification: Relevance toAdhesion, VSP, Vol. 3, pp. 263–283.

6. Conclusion
The tensile strength of the prepared composites
composite is 150 MPa, bending strength is 147
MPa and impact strength is 68 kJm-2 and It will bestable up to 50 years. This jute-based polymer
composite shows extra-ordinary features which arevery promising to make jute based polymer
composite as a effective alternative of metallic orplastic materials. It is rust proof, saline resistant,lightweight, heat resistant, sound proof,
environmental friendly, Very low thermal
expansion, and damaged area can be sealed very
easily. In contrast, lightweight jute made composite
boards fixed on steel frames with bolts & nuts are
more flexible allowing lateral movements of thestructures. They absorb and reduce seismic energy.
The usage of natural fiber (jute) based products inpost disaster management of rehabilitation &rebuilding, would become cost competitive
compared to other building materials. Thus, all theproperties of composite claim its position as an
ideal building material.
REFERENCES[1] Soumitra Biswas, Atul Mittal & G Srikanth,
Gujarat Earthquake - Composite Materials towards
Re-building & Rehabilitation[2] Bogoeva-Gaceva G., Avella M., Malinconico
M., Buzarovska A., Grozdanov A., Gentile G. andErrico M.E. (2007). Natural fiber ecocomposites.Polymer Composite;28: 98-107.
[3] Joshi S.V., Drzal L.T., MohantyA.K.andArora
S. (2004). Are natural fiber composites
environmentally superior to glass fiber reinforced
composites? Composites:Part A; 35: 371-376.
[4] Khan M.A., Ali K.M. I., Balo S.K. and Ahmad,
M.U. (1998). Effect of additives reinforcement ofradiation induced jute urethane polymer composites.
Journal of Applied Polymer Science; 67:79-85.

244

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Evaluation of Strength of Granulated Construction Waste Sludge with
Quicklime and its Application to Soil Structure
Makoto Kobayashi, Department of Civil Engineering, Toyota National College of Technology,
Hiroaki Nakamura and Hideaki Mizuno, Taihei Sangyo Co., Ltd

ABSTRACT: In this study, the strength characteristics of granulated
construction waste sludge mixed with quicklime were examined to
identify an efficient improvement method for the effective use ofsuch waste sludge. The tests results indicate that it is possible toestimate the strengths of improved sludge based on mixing ratio ofquicklime and dried sludge. Furthermore, appropriate curing isnecessary when using the sludge because the increased strength of
the improved soil depends on minimal water content, which caneasily increase due to saturation from rainfall. An in situ
investigation into applications of soil structures using improved soil
are also described in this paper, specifically, the counterweight fill
for preventing slope failure and forming dike for a reservoir.

Keywords: granulated construction waste sludge, quicklime,
soil structure, unconfined compression strength
1. INTRODUCTION
According to the white paper on Land, Infrastructure,
Transport and Tourism in Japan1), construction waste should
be reduced, recycled and reused to establish a soundmaterial-cycle for our society. The recycling rate ofconstruction waste sludge (CS) has been lower than that ofother construction waste. The primary reasons for thisproblem are as follows2), 3): 1)the characteristics of CS are
extremely complicated because it is generated by the additionof water, regardless of the type of soil used, and 2)thetreatment cost of CS is higher than that of construction
generated soil, since it is classified as industrial waste. It isnecessary to find an appropriate improvement method to raise
the recycling rate of waste sludge.
The properties and performance of CS treated with cement
and lime have been studied4), 5). With cement treated soil,
hexavalent chromium is generated, negating some of theadvantages of the resulting high strength. With thisshortcoming in mind, the characteristics of the soil stabilizedby lime should be investigated to use it as an alternative tocement treated soil. CS treated with quicklime is known forbeing slow in developing an acceptable strength. However,
the improved sludge should maintain the cone index just after
improvement at a value above 800kN/m2 that classified thesecond type of improved soil. To identify the amount ofquicklime that provides the most improvement in the CS, theunconfined compression strength should be tested after
quicklime are mixed and cured.
The improved CS is being considered for use as constructing
embankments and back filling6).Therefore, it is important toexamine the variation in the strength of the improved soil
when the degree of saturation is increased by rainfall.

In this study, to evaluate the performance of granulated
construction waste sludge mixed with quicklime (improvedsoil), cone penetrating tests and unconfined compression tests
were conducted. The improved soils were produced with CS
from various construction sites mixed with different
quantities of quicklime. The relationships between the cone
index and the unconfined compression strength just after
improvement were identified. Unconfined compression tests
were carried out to investigate the strength characteristics
increase resulting from curing. Based on these tests, a methodof describing the unconfined compression strength wasdeveloped based on the ratio of quicklime and dried sludge.
Furthermore, laboratory tests and in situ test were conducted
to examine the variation in the strength of the improved soil
when the degree of saturation was increased. Additionally,
the use of improved soil to form counterweight fill to prevent
slope failure and to construct a reservoir dike were observed
and discussed.
2. GRANULATED CONSTRUCTION WASTESLUDGE WITH QUICKLIME
2.1 Improvement technology

Fig. 1. The plant of treatment
The treatment plant is shown in Fig. 1. The improvement
method was as follows: 1) the CS collected in the pit was
dredged by backhoe and put on the belt conveyor to thetreatment plant. 2) On the belt conveyor, the CS was vibrated
and the quantity was averaged by a gate in the plant. 3) Then,the quicklime was added to the CS and mixed by a paddle
mixer. The amount of quicklime added was determined basedon the water content of the CS. No tests were done to identify
the CS properties before improvement. This choice was madeto increase the efficiency of improved soil production. On the245Unconfined compression
strength，qu (kN/m2)

2.2 Specimen
Improved soil was allowed to cure for 1day, and then, asample was compacted into a mold. This was followed bycarrying out the method for determining the cone index ofcompacted soils: JIS A 1228:2000. The specimen created forthe unconfined compression test was formed into a cylinder
10cm high and 5cm in diameter. It was then wrapped in vinyl.
Curing was done following one of two methods in this study.
The first method was standard curing while wrapped. The
second method was to cure the sample without wrap, natural
drying, in order to allow the water content to decrease withtime. Table 1. lists the test cases in this study, and for eachcase it lists the test code, the weight of quicklime relative tothe volume of CS, the initial water content of the improved
soil, and the weather conditions at improvement.
The permeability of the improved soil for test case CS38 was
k=9.3×10-7(cm/sec). This implies that it should be difficult
to saturate a soil structure constructed using improved soils.

strength and cone index just after improvement
The cone penetration test conducted to identify the quality of
the improved soil. Accordingly, it was also important toinvestigate the relationships between the cone index and theunconfined compression strength just after treatment. This
relationship is shown in Fig. 2.
For test cases with a lower cone index, the compression
strength was represented by. The. This result
other cases can be expressed by
implies the cone index will be representative of theunconfined compression strength7).
For test cases with high water content, a smaller unconfinedcompression strength was obtained. This was due to overcompaction during the compaction process. This is why two
versions of the estimation formula were required.
4 INCREASING STRENGTH WITH CURING
4.1 Standard curing

400
Compression stress，σ (kN/m2)

other hands, not reviewing the properties of the CS before
improvement makes it
difficult to obtain consistent
properties in the produced improved soils.

150

100

qu=0.13qc

qu=0.18qc

qu=0.1qc
50

Just after
improvement
1day curing
5days curing
7days curing
14days curing
28days curing

300
200
100
0
0

200

400
600
800
1000
Cone index，qc (kN/m2)
Fig. 2. Relationships between unconfined compression
strength and cone index shortly after soil improvement

3 UNCONFINED COMPRESSION STRENGTH
3.1 Relationships between unconfined compression

510Strain，ε (%)
Fig. 3. Stress-strain curve for test case CS18 with standard
curing
The stress-strain curves obtained from the unconfinedcompression tests for test cases CS18 and CS30-1 are shownin Figs.3 and 4, respectively. The longer the curing wasallowed to occur, the smaller the strain at the peak strength.
This implies that the improved soil will become a brittle
material after lime stabilization. The modulus of deformation

Table 1. Test code and condition of specimen as well as weather conditions
Weather condition at improvement

Test code

Weight of quicklime against
the volume of CS (kg/m3)

Initial water content of
improved soil(%)

Amount of precipitation(mm)

Average daily temperature(degree centigrade)

CS18

18

28.6

3.0

20.7

CS24

24

28.5

24.5

16.6

CS30-1

30

28.3

0.0

19.5

CS30-2

30

36.2

0.0

14.0

CS38

38

41.2

0.0

12.0

CS44

44

38.2

0.0

8.4

246

for test case CS18 after 28 days was 10MN/m2, showing an
increase since just after improvement the value was
2.5MN/m2. In test case CS30-1, the modulus of deformation
also increased by 4 time after quicklime stabilization.

Just after
improvement
1day curing
5days curing
7days curing
14days curing
28days curing

100

(1)

50
Water content，w (%)

Compression stress，σ (kN/m2)

200

quicklime to dried CS (Lime content: LC) was defined,
because it could be regulated in the treatment plant witheasily additional soil tests. The increment ratio of strength
was assumed by means of the relationships shown in Fig. 7. In
this figure, the rough relationships between increment ratio ofstrength and mixing ratio of quicklime were obtained asfollows:00510Strain，ε (%)
Fig. 4. Stress-strain curve for test case CS30-1 with standard
curingUnconfined compression
strength，qu (kN/m2)

300

10
20
Curing days (day)30

50

10 qu 28/qu ini = 10(0.33*Lc) - 0.8
5
qu 28/qu ini = 10(0.25*Lc) - 0.5

1
1

2
3
4
Mixing ratio of lime and CS，Lc(％)

Fig. 7. Relationships between increment ratio of strength andlime content

200
100
0
0

30

Fig. 6. Variation into water content vs curing period

400
CS18
CS24
CS30-1
CS30-2
CS38

CS24
CS30-240

20
0Increment ratio of strengthqu28/qu ini

The change in the unconfined compression strength inrelation to the curing period for in each test case in this study
is shown in Fig. 5. It is clear that the improvement in the soil
strength from the addition of the quicklime was slow-acting,so that the curing continued a full 84 days after improvement.
The variations in the water content of the specimens during
the curing process are shown in Fig. 6. A steady decrease in
the water content over time was not indicated in spite of thefact that the pore water should be spent not only on hydration
but also the pozzolanic reaction of the improvement
mechanism. As can be seen from the Fig.5, after 28days ofcuring the strength of the improved soil doubled. However,the figure also implies that a small amount of quicklime isenough to improve the CS.

CS18
CS30-1
CS38

20

40
60
80
Curing days (day)Fig. 5. Variation in the unconfined compression strength
relative to curing period.

In order to analyze the unconfined compression strength of
improved soil after 28 days curing (qu28), first an increment
ratio relating qu28 to qu ini was defined. The term qu ini indicatesthe unconfined compression strength just after improvement,
as identified using the one index. Second, the mixing ratio

It is difficult to manage the lime content relative to several
properties of the CS while the treatment plant is producing
improved soil at a rate of more than 500m3/day from a diverse
selection of construction sites. As a result of these tests,
however, the strength of the improved soil at 28 days can beroughly estimated using the cone index, lime content, andwater content of CS.
4.2 Curing with natural drying
The stress-strain curves obtained by the unconfinedcompression tests are shown in Fig. 8. The peak strength after
curing for 5 days was 8 times the initial strength. These
specimens suffered brittle fracture after peak strength wasachieved.

247

Compression stress，2
σ (kN/m )Just after
improvement
1day curing
5days curing
7days curing
14days curing
28days curing

600
400

Water content，w (%)

the permeability of these materials is lower than that ofgeneral soil when used for soil structures, so only the surfaceof the structure made with improved soil will become
unstable due to rainfall.

800

200
0
0

Unconfined compression
2
strength，qu (kN/m )The relationships between the unconfined compression
strength and the length of the curing time are shown in Fig. 9.
In this figure, not only is the increment of strength observable,
but also the sudden decrement of strength in the cases of
CS30-2 and CS38. Fig. 10 shows the change in the watercontent of the specimens during the curing process. In all test
cases, the water content decreased until about the tenth day
before stabilizing. This indicates that the changes in thecompressive strength were not caused by lime stabilization
but rather by increase the friction between soil particles thatcomes with a decrease in the water content.

30
20
10
0
0

10
20
30
Curing days (day)Fig. 10. Variation in water content with curing

Compression stress，2
σ (kN/m )5
10Strain，ε (%)
Fig. 8. Stress-strain curve for test case with CS30-1 for curingwith natural drying

CS18
CS24
CS30-1
CS30-2
CS38

40

300
200

Curing for natural drying
28 days curing
Submerged

100
0
0

800

Standard curing
28 days curing
Submerged

24Strain，ε (%)

6Fig. 11. Stress-strain curve when the sample is submerged

600

5 APPLICATION TO SOIL STRUCTURE

400
200

CS18
CS30-1
CS38

CS24
CS30-2

0
0

10
20
30
Curing days (day)Fig. 9. Variation in unconfined compression strength relative
to curing time

If the above mentioned theory is correct, it is important toinvestigate the unstable mechanism during rainfall, which
will increase the volume of void water. To test this condition,
samples were submerged in water before compression testing
to test the strength when saturated. Using a specimen (CS38)
that had cured for 28 days, the unconfined compression test
was conducted after the specimen had been submerged for
24hours. The variation in the stress-strain curve obtained bythis test is shown in Fig. 11. The effect of submergence was amarked decrease in the peak strength when the sample wascured by natural drying. This was due to the increased watercontent lubricating the motion of the soil particles. Namely, it
implies that the compression strength of improved soil cured
by natural drying increases with the increase of friction of thesurface, which dependent on the moisture levels. However,

5.1 The counter weight fill for slope failure
Cracks in the slope surface were observed at a ceramicmaterial mine where the site had been reclaimed by using
construction generated soil, as shown in Fig. 12. Identifying
the cause of this phenomenon was complicated because of thefollowing: 1) ground water was existent in the slope and 2)
the load on the slope was increased due to the construction ofan embankment at the crest of the slope.
To solve these problems, improved soil was used to form acounterweight fill at this site. Also, the slope angle waschanged to a gentler 1:2.0, or 26.6 degree. Lastly, a drainage
conduit was inserted into the slope in order to lower the
ground water level.

248

(a) Analysis model at sliding(b) Analysis model of countermeasure
Fig. 13. Analysis model of the groundFig. 12. Observed cracks in slope
To examine the effectiveness of the countermeasure to thelandslide, FEM analysis was performed. In this analysis, theshear strength reduction method (SSRM)8) was used toinvestigate the critical slip circle of the slope. The analysis
model and the parameters of the ground material are shown in
the Fig. 13 and Table 2, respectively. The results ofunconfined compression tests with CS30-1, which examined
the long term strength, were used to estimate the cohesion of
the improved soils. In addition, the result of a triaxial
compression tests was used to identify the internal friction
angle of the improved soil. The internal friction angle of
Improved Soil 1 was used in the model, with a measure ofunderestimation to account for the ground water present in theslope. Other parameters for the ground materials were
decided by analysis, which continued until similar
deformation was obtained.
The results of the FEM analysis, which expressed the
distribution of the maximum strain increment obtained by
SSRM, are shown in Fig. 14. The critical slip surface wasconfirmed through the upper side of the embankment madeby the construction generated soil and down to the toe area ofslope. Furthermore, the largest strain was generated in thebase ground at the toe area of the embankment. This indicatesthat the crack was caused by differential settlement of thefoundation. After countermeasures were installed using
improved soil, however, there was no definite slip circle inthe slope.

(a) Result of analysis at slide(b) Result of analysis after countermeasure
Fig. 14. Distribution of maximum strain
5.2 Dike of regulating reservoir
The improved soil was also suitable for constructing a dikefor a regulating reservoir because of its low permeability, as
already mentioned in 2.2. The photo of dike underconstruction is shown in Fig. 15. In the 2 years sinceconstruction, no remarkable deformation has been observed.
Further observation is continuing, watching for changes,
including deformation and collapse.

Table 2. Parameter of ground material for FEM analysis
Material No.

Model

Young’s modulus(kN/m2)

Cohesion(kN/m2)

Internal friction angle(degree)1

Base ground

2×105

-

-

2

Construction generated soil

8×102

1.0

24

3

Improved soil 1

1.5×104

30.0

20

4

Improved soil 2(for countermeasure)

1.5×104

200.0

24

5

Base ground under toe of slope

4×102

1.0

20

249

Constitutive law
Elastic
Elastic-perfectly plastic
Mohr-Coulomb
Elastic-perfectly plastic
Mohr-Coulomb
Elastic-perfectly plastic
Mohr-Coulomb
Elastic-perfectly plastic
Mohr-Coulomb

[8]Fig. 15. Dike of the regulating reservoir during construction

6 CONCLUSION
In this study, a series of soil test were carried out to estimate
the performance of granulated construction waste sludge
when mixed with quicklime. Furthermore, a case study of the
improved soil when used in two soil structures was described.
The conclusions of this study as follows:
1) The estimated strengths of improved soil can be estimatedbased on the mixing ratio of lime to dried sludge and the cone
index of the improved soil just after treatment.
2) The improved soil must be cured to fully strengthen thesludge since the increased strength of improved soil depends
on a low water content and can easily be reduced withsaturation by rainfall.
3) The improved waste sludge is applicable for some soil
structure, for instance, as counterweight fill to prevent slope
failure and forming a dike for a reservoir. The granulated
construction waste sludge has a low permeability and large
strength while being easily formable.
7 REFERENCE
[1]
[2]

[3]

[4]

[5]

[6]

[7]

Ministry of land, infrastructure, transport and tourism(2009): White
paper on land, infrastructure, transport and tourism in Japan, pp. 93-94.
Kamon, M. (2008):Current status and recycling strat-egies of
construction sludge, Waste Management Research, Vol.19, No.3,
pp.122-130.
Sakamoto, H. (2008): Approaches to construction sludge recycling in
the construction industry, Waste Management Research, Vol.19, No.3,
pp.138-145.
Ogawa, N., Akimoto, K, Seki, S. and Yoshinari, S., (1995): An attempton effective use of construction sludge (part1), The 30th japan national
conference on soil mechanics and foundation engineering,
pp.2221-2222.
Umemoto, K., Kirikoshi, S., Fujisaki, K. and Ohhara, T., (1995): An
attempt on effective use of construction sludge (part3), The 30th japan
national conference on soil mechanics and foundation engineering,
pp.2225-2226.
Onodera, Y., Yoshida, H., Tajika, H., Miyakawa, M. and Saito, K.
(1999): Mechanical properties of stabilized gravelly soft soil, -Backfill
mate-rials of pipeline with stabilized soil by lime-, Monthly report ofcivil engineering research in-stitute, No.556, pp.32-41.
Ogawa, N, Kirikoshi, N, Hisano, K. and Yamamoto, H., (1995): An
attempt on effective use of construction sludge (part4), The 30th japan

250

national conference on soil mechanics and foundation engineering,
pp.2227-2228.
Ukai, K., (1990): Availability of shear strength reduc-tion method instability analysis, Tsuti-to-kiso JSSMFE, Vol. 1, No.38. pp.67-72.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Analysis of Enhanced Strength and Deformation Resistance of Some Tropical
Geomaterials through Application of In-situ Based Stabilization Techniques
Kogi SK & Ndeda M., Materials Testing & Research Department, Ministry of Roads, Nairobi, Kenya
Mukabi JN & Wekesa S., Research & Development Department, Kensetsu Kaihatsu Limited, Nairobi, Kenya



ABSTRACT: In this paper, some aspects of OBRM (Optimum Batching
Ratio Method) and OPMCS (Optimum Mechanical & Chemical
Stabilization) in-situ based stabilization techniques, which have beenscientifically developed and successfully applied for the design andconstruction of runway and highway pavement structures in East and Central
Africa, are introduced through experimental data analysis.
Quantitative analyses are carried out to show how the improved mechanicalstabilization and reduction in voids (densification) by accelerated particle
agglomeration, through the application of OBRM and OPMC respectively,
contribute significantly to the enhancement of bearing capacity, strength anddeformation resistance of the composite geomaterials.
It is then demonstrated, through comparative analysis of the enhanced
geotechnical engineering properties in respect to the cost aspect ratio andenhanced geotechnical engineering properties, that these techniques arehighly cost-time-maintenance-effective whilst contributing to environmental
conservation due mainly to the reduction in the quantities required forconstruction and length of access roads to borrow pits in comparison to theconventional and traditional methods of design and construction.

Keywords: OBRM, OPMCS, Geomaterial, Strength anddeformation resistance, Design and construction
1. INTRODUCTION
1.1 Geomaterial Constraints in Design and Construction
In most parts of the world and, developing countries inparticular, engineers are constantly faced with the major
challenges of designing, constructing and maintaining
pavement structures with limited financial resources. On theother hand, for the fulfillment of the acceptable construction
standards, traditional design and construction practices
require high-quality geomaterials which are usuallyexpensive and mostly involve long haulage distances thereby
increasing the period of construction and, most of all, theproject cost. Due to these constraints, engineers are often
forced to seek alternative designs that are innovative enoughto allow for the use of substandard geomaterials, which, insingularity, are mostly deficient in mechanical stability andother geo-properties but are, however, readily available forthe construction of most layers of the pavement and othergeo-structures such as embankments and bridge abutments.
1.2 Brief Introduction of OBRM and OPMCS
Technologies
In their natural state, most geomaterials are usually deficientin one or more of the particle fractions required.
Consequently, mechanical stabilization plays an importantrole in achieving a pavement structure which, under loading
conditions, is appreciably resistant to shear and deformation.
In developing the Optimum Batching Ratio Method (OBRM),[1] considered that; such geomaterials would have a particle

size distribution that tends towards a correctly proportioned
ratio that would yield optimum density and adequate strength
to resist stress-induced deformation.
(1) Optimum Batching Ratio Method (OBRM)
The OBRM is a mechanical stabilization method that wasodeveloped on geo-scientific basis. Based on this method, theblending of two otherwise structurally inferior geomaterials
with different characteristics can be achieved at optimum
proportionate levels thereby enhancing the vital geotechnicalengineering properties such as mechanical stability, bearing
capacity, strength and deformation resistance [1], [2].
(2) Optimum Mechanical and Chemical Stabilization(OPMCS)
OPMC is the combination of optimum mechanical batching(OBRM) and optimum chemical mixing of, and into
geomaterials with varying properties [3], [4].
These methods have been practically applied for the designand construction of various civil engineering projects in East
and Central Africa and have been proven to: 1) promote
cost-effective use of otherwise relatively inferior
geomaterials, whilst ensuring the achievement of soundgeotechnical engineering structures; 2) contribute immensely
to environmental conservation through the reduction in use ofborrow and/or extruded geomaterials; 3) provide partial
solutions to problematic soils, and; 4) realize significant
reduction in the requirements of chemical stabilizers.
2. METHODS OF TESTING AND ANALYSIS
2.1 Laboratory and In-situ Testing Testing
Details of the standard and innovative laboratory and in-situ
testing regimes and methods are discussed in [5]-[9].
Experimental testing was undertaken on soil samples from,and in-situ, in five different countries in East and Central
Africa namely; Southern Sudan, Ethiopia, Burundi, Tanzania
and Kenya.
2.2 Full Scale Experimental Testing Sections

Full-fledged field experimental sections with varying
pavement structural configurations were designed andimplemented as reported in [6]-[10].
2.3 Main Analytical Functions Adopted
The fundamental theories and concepts adopted for thecomprehensive analysis are discussed in detail in [4], [10],whilst some of the basic functions are introduced in thissection. The analytical functions were derived to characterize251the impact of environmental factors on the performance of theroad pavement, evaluate the change in the intrinsic material
characteristics influenced by the nature, mode and degree ofstabilization coupled with the reciprocal impact and intensityof loading [4] and [10], [11].
The relations applied for comprehensive Quality Control(QC) of OBRM and OPMC stabilized ground andgeomaterials are also proposed in [11].
(1) Analysis of Mechanical Stability, Bearing Capacity,
Shear Strength and Deformation Resistance
The analyses for mechanical stability, strength anddeformation resistance were made based on various equations[1]–[4] and [10]. Some of the principle relations adopted in
the analysis of this Study are presented in (1) ~ (11).
Quantitative analysis of the contribution of enhanced
mechanical stability due to OBRM stabilization wasundertaken by applying (1) to (5).

M

f
S

 A MS exp

when η ≥ 0.5

 1.16

(1)



M Sf  M Sideal  AMS exp1.16  M Sideal

when, η>0.5

BC  B

ideal
C

(2)



 ABC  AGI  B



(8)

Bf

where, (qu)max values are expressed in kN/m2 and ANf, Af, Bf
are experimentally determined geomaterial constants.
On the other hand, under triaxial conditions considering thatq = ( ' a  ' r ) max and P ' f  1 3  ' a  2 ' r  then,




q max

 2 p ' f  1 3 q max  '  Sin 1 

(9)The elastic stiffnessof OPMC stabilized Geomaterial
taking into account the particle agglomeration effects, isquantitatively analyzed from: a ij50
   ij
ELS  a ijmax  A, (%)
ij
a ELS

(4)

when, 0 < η < 0.5f

B C   exp A BN M S

ANf (qu )max  Af

62.4
690 are geomaterial
where,
constants.
is measured and/or expressed in MPa.
Estimation of the linear elastic range or initial yield surface ismade from (11).

(3)
ideal
C

' 

(10)

when 0 < η < 0.5

B C  ABC  AGI 

where,
0.05 and
0.95 are geomaterial constants.
is determined
Based on UCS testing of OPMC specimens,from (8).

(5)

where, M is the Mechanical Stability (MS) factor, and forthis case, AMS is the MS constant=178.6, η is the gradationindex=log0.01P/log(d/dmax), MSideal=100% for non-stabilizedgeomaterial, BC is the bearing capacity factor, ABC is the BC
constant=130, AGI is the grading index constant=60,
BCideal=100% for non-stabilized geomaterial, δ=BC-MS
correlating constant=24.7 and ABN =BC-MS constant=0.014.
The contribution of particle agglomeration in enhancing
strength and elastic modulus due to cementation isquantitatively analyzed from (6).
(6)
where,
is the maximum shear strength,1.06 is amaterial constant andis the agglomeration factor
computed from the level of OPMC related factors.
On the other hand, the effective angle of shearing resistanceis computed as:
(7)

(11)

where, ϕELS is a function of the deformation modulus E50 withand A is a
respect to the initial (Elastic) modulus,
constant depending on the physical properties of thegeomaterial.
3. ANALYSISRESULTSANDDISCUSSIONS

OF

TEST

3.1 Tabulated Summary of Some Typical Test Results
Tables 1 and 2 present a tabulated summary of some typicalresults based on laboratory and in-situ experimental testing
reported in this Study. Table 1 is a summary of the basic
physical and mechanical parameters for non-stabilizedgeomaterials, while part of the measured shear strength andelastic modulus data is presented in Table 2.
The results in table 1 show that: 1) the basic physical andmechanical lateritic soils from different countries vary
drastically (the same was observed for lateritic materials
sampled from different locations in the same country); 2) the
relatively fine graded Kenyan lateritic gravel exhibits rather
superior bearing capacity, shear and deformation resistanceproperties when compared to the relatively coarse graded one
from Burundi; attributable to the superior mechanicalstability of the former [10]; and, 3) the degree of influenceof MDD and OMC depends on properties of the geomaterial.

252

Table 1 Summary of basic physical and mechanical geo-properties of non-stabilized geomaterialType of material

Compaction
Characteristics
MDD
(Kgf/
cm3)

Southern
Sudan

Jonglei

Juba

Ethiopia

Addis
Addis
Addis

BurundiTanzania
Kenya

Bujum
-bura
Songw
e
IsioloConsolidation

Lateritic Quartzic
Wearth. Gravel
Sandy Clay
HighlyWeathered
Clayey Gravel
Volcanic Ash
Graded Crushed
Basaltic Ag.
Coarse Graded
Lateritic Gravel
Pozzolanic
Fine Graded
Lateritic Gravel

OM
C
(%)

Densification /Atterburg Properties

FM
C
(%)

VR
e

GS

PI
(%)

Clay
Acti
vity
AC

MS
(%)

Bearing

Basic Physical Parameters

Shear
Properties

Capacity

Geomaterial

Mec. Stab.

Location

Elastic
Modul
us

′

CBR
˚

Sieve

CountryofOrigin

UCS
/CU
TC

UCS/
CU
TC

CBR/
DCP

UCS/
DCP

CU
TC

CUTC/
UCS
DCP

2141

7.4

6.0

0.17

2.75

17.5

0.42

14.2

46.1

1.565

30

0.71

26.7

392

1650

19.6

32.8

0.86

2.62

25

0.64

-18

28.1

1.494

2.6

0.087

23.6

18.8

1640

20.3

11.9

0.42

2.21

18

0.46

-12

47.6

1.569

21

0.51

28.3

345

1678
2322

16.7
7.8

9.9
5.7

0.38
0.18

2.13
2.73

8
0

0.28
0

71.2
105

48.2
57.3

1.571
1.594

67
108

1.62
2.62

29.5
37.3

1883
3403

2120

6.9

12.5

0.35

2.79

10

0.42

32.7

41.7

1.551

39

0.94

27.8

478

1780

14.5

12.6

0.26

2.03

6.4

0.16

61

48.3

1.571

58

1.40

29.8

1467

1486

19.1

15.3

0.36

2.36

11

0.53

53.3

45.3

1.562

52

1.23

28.6

1216

Table 2 Summary of shear strength and elastic modulus of OBRM and OPMC stabilized geomaterials
CountryofOriginLocation

Type ofmaterial

Juba

Lateritic
Quartzic
Wearth.
GravelSandy Clay
HighlyWeathered
ClayeyGravelVolcanic
Ash
GradedCrushedBasaltic Ag.
CoarseGradedLateritic
GravelPozzolanic

Southern
Sudan
Jonglei
Addis

Ethiopia

Addis

Addis

BurundiTanzania
Kenya

Bujumbura
SongweIsioloFine Graded
Lateritic
Gravel

OBRM1

1.21

1187

OBRM2

1.93

OPMC1

3029

2.66

3422

OPMC2

3.11

5201

OPMC3

4.26

5861

OPMC6

8.13

7492

0.27

125

0.52

301

1.28

1244

1.66

1880

2.59

3388

2.83

3504

0.63

669

0.96

681

1.71

1568

2.21

3190

2.72

3452

3.35

3736

1.86

2823

2.35

3085

2.72

4942

3.36

5356

4.75

6108

6.13

6730

3.21

3676

3.96

3981

4.93

6195

5.95

6654

7.65

7321

9.88

8069

1.35

1923

1.79

2356

2.26

4606

2.64

4886

3.12

5207

6.48

6874

2.86

3518

4.67

4239

6.11

6721

7.25

7173

8.69

7685

11.74

8615

1.74

2913

2.30

3239

2.97

5110

3.87

5651

4.71

6089

5.63

6516

It can further be noted from Table 1 that the Mechanical
Stability (MS) is higher for the Kenyan gravel than the
Burundi one. It can therefore be inferred that MS is thecontributing factor to the enhanced shear and deformation
resistance properties. A graphical summary of part of theresults presented in Table 2 is made in Fig. 3.4.
3.2 Shear Strength(1) OBRM Characteristics
A typical graphical representation of OBRM shear strengthcharacteristics is presented in Fig. 3.1.

Fig. 3.1. Graphical representation of OBRM shear strengthcharacteristics of geomaterials sampled from the swampy
Sudd Plains in Jonglei, Southern Sudan.

In this case, the Optimum Batching Ratio (OBR) was 70:30 inpercentage terms of relatively expansive Sandy Clay : Sand.
The original sand content in the sandy clay wasapproximately 12%. This exercise was part of the Basic
Design (BD) study for pad foundations for oil drilling rigs.
It can be noted from this figure that, as the batching
proportionate tends towards to an optimum value, the shear
strength is enhanced accordingly notwithstanding the inferior
quality of the base material.
(2) OPMCS Characteristics
The OPMCS shear characteristics for specimens subjected tovarying modes of stabilization, soaking and curing conditions
are depicted in Fig. 3.2. The testing regime was specificallydesigned to investigate the rate of particle agglomeration fordifferent batching ratios and types of chemical agents.
OPMC-co implies the geomaterials that were batched atratios of OBRM, i.e., 50:25:25; Lateritic-Quartzic Gravel :Coarse Aggregate : Fine Aggregate with an addition of 20%
by weight of Portland Cement (PC), while OPMC-cl refers tosimilar OBRM ratios but with an addition of 14% PC and 6%
Hydrated Lime (HL).
Analysis of the OPMC stabilized Juba gravel indicates thatthe following derivations can be made: 1) the development of
most strength and deformation resistance is realized within253

countries in East and Central Africa is presented in Fig. 3.4.
Shear Strength, UCS (MPa)the first three days (also ref. to Fig. 3.3); 2) It can be noted
that, in this case, the OPMC-cl attains the commonly
specified strength of qu ≥ 30kgf/cm2 after only three days ofcuring while the OPMC-co will have attained 65% of therequired strength within the same period, and; 3) the inclusionof lime seems to accelerate the particle agglomeration
mechanism by a significant magnitude. This mode ofstabilization can therefore be effectively applied forgeo-structures that require enhanced development of bearing
capacity, strength and deformation resistance within a shortperiod of time hence reducing the construction time.

14.00

Juba Lateritic‐Quartzic Gravel

12.00

Jonglei Sandy Clay
Addis Clayey Gravel

10.00

Addis Volcanic Ash

8.00

Addis Graded Crushed Aggregate

6.00
4.00

Bujumbura Coarse Gravel
Songwe Pozzolana
Isiolo Fine Graded Gravel

2.00
0.00

Neat OBRM1 OBRM2 OPMC1 OPMC2 OPMC3 OPMC6
Mode of Stabilization

3.3 Initial Deformation Resistance (Elastic Modulus)
The investigation results on the influence of curing period in
relation to the initial deformation resistance are graphicallyplotted in Fig. 3.3 for the two modes of stabilization
introduced in this Study. It can be observed that: 1) as thecuring period prolongs, the development of the elastic
modulus mildly tends towards a residual state; 2) in general,
the rate of strength and deformation resistance is higher for
the OPMC-cl specimens compared to the OPMC-co ones;
and, 3) the addition of lime enhances the rate of developmentof the deformation resistance immensely.

3.5 Combination of OBRM/OPMCS and Geogrid
Reinforcement
The quantitatively analyzed effects of soil particle
agglomeration on the Unconfined Compressive Strength(UCS) for varying modes of stabilization are depicted in
Figure 3.5. The results were adopted in the Detailed Design(DD) of the runway pavement, aprons, taxiway and parking
for the Isiolo Airport located in the North Eastern Province of
Kenya, which is to be upgraded to international standards [8].
Agglomeration UCS (MPa)Figure 3.2 OPMC-cl and OPMC-co Stress – strain relations
for varying curing and soaking conditions for specimens
constituent of geomaterials sampled in Juba, Southern Sudan.

Figure 3.4 Summary of shear strength enhancement based on
different modes of stabilization for varying tropicalgeomatrials
It can be derived from Fig. 3.4 that: 1) response to the modeof stabilization depends on the intrinsic properties of thegeomaterial; 2) the magnitude and rate of increase in the shear
strength varies with the mode of stabilization; 3) the rates ofincrease in the magnitude of shear strength are virtually
uniform for the Addis Graded Crushed Aggregate andSongwe Pozzolana; and, 4) shear strength enhancement isdependant upon the degree/level and mode of stabilization.

4.50
4.00
3.50
3.00
2.50
2.00
1.50
1.00
0.50
0.00

7 Days Cure/Soak
3 Days Cure/Soak
28 Days Cure/Soak

Neat

Figure 3.3 Effect of curing period on the elastic modulus of
OPMCS geomaterials – Juba, Southern Sudan.
3.4 Graphical Summary of OBRM/OPMCS Shear
Characteristics
The graphical summary of OBRM/OPMCS shear
characteristics subjected to different modes of stabilizationfor varying tropical geomaterials sampled from a number of

OBRM
OPMC OPMC + Geogrid
Mode of Stabilization

Figure 3.5 Enhanced strength resulting from particle
agglomeration for varying modes of stabilization [12].
It can be noted from this figure that: 1) the UCS issignificantly enhanced as the soil particles agglomerate
progressively with time particularly for the cemented
geomaterial, 2) soil particle agglomeration is dependent on
the mode of stabilization, 3) the effects of geogrid
reinforcement are more noticeable with increased
agglomeration possibly due to the influence of confinement
in increasing the degree of interlocking of particles
culminating in reduction of voids ratio and, 4) the degree ofsoil particle agglomeration is higher under OPMC
stabilization in comparison to geogrid reinforcement thus
further confirming the theory that, in comparison to other254

influencing factors, cementation processes have greater
impact on agglomeration mechanisms .

5. EVALUATIONOFCONSTRUCTION
MAINTENANCE REQUIREMENT RATIOS

4.

The comparative evaluation and analysis of the construction
and maintenance requirement ratios expressed in terms ofenhanced structural soundness, cost and time savings, wasundertaken for the recently constructed Songwe International
Airport located in the Southern part of Tanzania [8].
The results of three varying design options are shown in Figs.
5.1 ~ 5.6. OPTION1 (Reviewed) is the design that adopted
the OBRM/OPMCS technologies reported in this Study.

CONTRIBUTIONOF
OBRM/OPMCSTOENVIRONMENTAL IMPACT MITIGATIONIn this Study (ref. to Fig. 4.1) as well as [6]-[10], the results
have shown that through the application of the OBRM andOPMCS technologies: 1) reduction of volume of materials
used by approximately 40% is achieved in most cases; 2) lessdisturbance of land for borrow pits; 3) reduced amounts ofdisposable soil during construction; 4) reduced risk ofcollapse of geo-engineering structures and; 5)environmentally friendly due to; utilization, as much as
possible, of locally available material and reduction of dust,
distances (lengths of access roads) to borrow pits, geomaterial
quantities required and land acquisition, among other factors.
Whilst developing these methods, comprehensive appraisals
and environmental assessments that would lead to sustainable
development with minimal negative environmental impacts,
were undertaken [11].
The example depicted in Fig. 4.1 was part of the design for
Wau~Abyei Trunk road constructed in the northern oil fieldsof Southern Sudan.

AND

5.1 Quantitative Analysis of Structural PerformanceFigs 5.1 and 5.2 clearly show that the OBRM/OPMCS
technologies significantly enhanced the structural capacityand deformation resistance of the pozzolanic geomaterials
exhibiting far more superior geo-properties of the composite
pavement structure in comparison to the other designs.

Figure 5.1 Quantitative derivation of OBRM/OPMCS
enhanced structural capacity performance.

Figure 5.2 OBRM/OPMCS enhanced initial deformation
resistance analysis in terms of elastic stiffness measured in
in-situ in MPa.
5.2 Cost Aspect Ratio for Construction
The cost aspect ratio was computed mainly on the basis of: 1)
reduction in the material quantities required; 2) reduction inconstruction time; 3) enhancement ratio of the structural andmechanical components in relation to the achievement of asound geo-structure, and; 4) maintenance requirement ratio.

Figure 4.1 Contribution of OPMCS to reduction ofenvironmental impacts of road works [11].
A summary of the approach, considerations and contributionof OBRM and OPMCS from an environmental andgeotechnical engineering perspective are discussed in [11].

Figure 5.3 Reduction in construction time due to applicationof OBRM/OPMCS technologies.

255

As can be inferred from the results depicted in Figs. 5.1~5.5,
application of OBRM/OPMCS ensures the achievement of
the stipulated objectives by an appreciable margin.

6 CONCLUSIONS

Figure 5.4 Significant reduction in construction cost as aresult of OPMCS (shown in blue: approx. 40% reduction).

7 ACKNOWLEDGEMENT

5.3 Maintenance Requirement Ratio (MRR)
The computation of the MRR was made from (12) based on
the results of the SCDR [8] model by adopting the TA over an
initial reference loading period Nt and Design Life,(DL=20years).

The analysis of the performance of some recently developed
in-situ based stabilization techniques have been reported inthis paper for varying tropical geomaterials from fivedifferent countries of the East and Central African Region.
It has been demonstrated in all cases that the OBRM/OPMCS
technologies contribute significantly to not only theenhancement of the geotechnical engineering properties thatare vital in design, but also in achieving immense
construction cost and time savings as well as minimum
maintenance requirements.

The authors wish to extend their sincere appreciation to theJapan International Cooperation Agency (JICA), Katahira &Engineers International, Japan and Urban Tone Corporation,
Japan, for sponsoring most of the research for this Study.
Acknowledgement is also extended to the Materials Testing& Research Department of the Ministry of Roads, Kenya, theTanzania Airports Authority, and the KGS for their support.
8 REFERENCES
[1]

.
.

.
.

(12)

Figure 5.5 Analytical results of the Maintenance Requirement
Ratio (MRR).
5.4 Summary of Overall Performance
A summary of the overall performance of the three pavement
structures based on the varying designs is presented in Fig.
5.6. It can be noted that OBRM/OPMCS excels in all cases.

Figure 5.6 Summary of performance of the varying designs
and technologies

Mukabi, J.N, “Theoretical and empirical basis for a method ofdetermining the optimum batching ratio for mechanical stabilization of
Geomaterials”, Procs. 14th IRF road World Congress, Paris, 2001, CD.
[2] Mukabi, J.N & Shimizu, N, “ Strength and deformation characteristics
of mechanically stabilized road construction materials based on a newbatching ratio method”, Procs. 14th IRF Road World Congress, Paris,
2001, CD-ROM.
[3] Mukabi JN, Kimura Y, Shimizu, N, Mwangi SN, Omollo A, Njoroge
BN, “ Evaluation of some Kenyan Geomaterial properties forembankment design based on a quasi-empirical approach”, Procs. 15th
Int. Conf. on SMGE, Istanbul, 2001, pp2159-2166.
[4] Mukabi JN, “Case Study Analysis of OPMC improved foundation
ground, pavements and geo-structures employing the GECPRO model”,
to be published in Procs. Of ISSMGE International Symposium on
Ground Improvement (IS-GI), Brussels, 2012.
[5] Mukabi JN, “Innovative laboratory and in-situ methods of testing ingeotechnical engineering, to be published in the Proc. of the Int. Conf.
of the Inst. of Engineers Kenya , Nairobi, 2013.
[6] Kensetsu Kaihatsu Limited, Juba River Port Access Road Pavement
Design, Engineering Report No. SST1, submitted to Japan International
cooperation Agency, JICA, vide Katahira & Eng. Int., Japan, 2007.
[7] Kensetsu Kaihatsu Limited, A Comprehensive Engineering Reporton the Study, Design and Construction of Drilling Pad Foundations in
Jale, Jonglei State, Southern Sudan, submitted to White Nile Oil
Exploration Company, London, England, 2007.
[8] Kensetsu Kaihatsu Limited, Songwe Airport Pavement Design Review
Engineering Report No. SAT 09T1, submitted to Tanzania Airports
Authority, Government of the Republic of Tanzania, 2011.
[9] Kensetsu Kaihatsu Limited, Isiolo Airport Pavement Design,
Engineering Report No. ISAT 0211/O2, submitted to Kenya Airports
Authority, Government of Kenya, Nairobi, 2011.
[10] Mukabi JN, Kimura Y, Murunga PA, Njoroge BN, Wambugu J, Sidai
V, Onacha K., Kotheki S, Ngigi A, “Case example of design andconstruction within problematic soils”, in Procs. Int. GeotechnicalConference on Geotechnical Challenges in Megacities, Geomos,
Moscow, 2010, vol. 2, pp 1172-1179.
[11] Mukabi J.N., Toda, T. & Shimizu, N, “Application of a new mechanicalstabilization technique in reducing the cost and impact of rural road
construction, Procs. 23rd World Road Congress, Paris, 2007, CD-ROM.
[12] Wekesa S, Mukabi JN, Sidai V, Kotheki S, Okado J, Ogallo J, Amoyo
G, Ngigi L, “Quantitative analysis to verify the theory of soil particle
agglomeration and its’ influence on strength and deformation resistanceof geomaterials”, Procs. 15th ARC on SMGE, Maputo, 2011, pp.
330-335.

256

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Further Developments in the Swelling Model of Expansive Clays under ASTM
4546 and CBR Testing
Moshe Livneh. Civil and Environmental Engineering Department, Technion-Israel Institute of Technology



ABSTRACT: The prediction of heave for pavements based on
expansive subgrades requires knowledge of the percentage
expansion expected at each depth at which the absorption of watertakes place until equilibrium moisture is obtained under a controlled
suction state. This heave prediction necessitates knowledge of theswell-pressure characteristics of the given clay stratum, data that areusually obtained from laboratory tests on undisturbed clay samples.
It has been demonstrated in regard to 352 local undisturbed samples
under ASTM 4546 testing that a general swelling model can be
employed that predicts the percentage of swell under a given verticalpressure exerted on clay characterized by basic parameters. The
present paper, too, deals with the development of the generalswelling model, but this time for 674 local CBR remolded samples
soaked in water and subjected to swell under a given surchargepressure for a period of 4 days. The paper concludes with acomparison of these two general models.

The rate of the remolded vertical swell ratio is determined,
obviously, by conducting specific vertical swell andswell-pressure tests on undisturbed and remolded samples fora given project. This paper, however, concentrates on thedevelopment of an additional swelling model for local CBR
remolded samples soaked in water and subjected to swellunder a given surcharge pressure for a period of four days.
This additional swelling model is based on the same
regression analysis used for the development of the generalswelling model referred to above. Finally the paper concludes
with a comparison of the these two models, one forundisturbed samples tested according to ASTM 4546, and thesecond for compacted samples in CBR molds after four daysof soaking in water.

Keywords: CBR, expansive-clay, heave, suction, swell.

2. OLD GENERAL SWELLING MODELS

1. INTRODUCTION

The first Israeli general swelling model was published in1969 [1]. This model introduces a statistical relationshipbetween the swell pressure (Po), in kPa, and the following
parameters characterizing the undisturbed clay samples:
liquid limit (LL) in percentages, moisture content (W) in
percentages, and dry density in kN/m3. The generalformulation of this model is as follows:

A considerable proportion of roads and railways are based on
swelling-clay subgrades. Thus, it is important to assess therates of heave expected to occur as a result of clay activity inany given alternative that defines the design solution forpavement or railway-bed structures. This assessment is basedon a swelling model that links the rate of vertical swell withthe clay’s characteristics, combined with the rate of the swellpressure (i.e., rate of the vertical pressure to obtain a zero rateof vertical swell).
The reliability of the aforementioned heave assessment
depends on, among other things, the reliability of the swellingmodel (also known as the general swelling model) used in theprediction. This swelling model refers to undisturbed samples
subjected to ASTM 4546 Method B testing for rates ofvertical swell and ASTM 4546 Method C testing for rates ofvertical swell pressure. Thus, it is only natural to have aninterest in updating this model with the accumulation of
additional test results over time. This paper presents such an
update, while comparing its output with previous results.
In addition, it is important to determine the rate of theremolded vertical swell ratio, defined as the ratio of thevertical swell rate obtained in remolded clayey samples to thevertical swell rate obtained in natural (undisturbed) clayey
samples. Obviously, this definition refers to equal conditionsof moisture, density, and consistency limits for both remolded
and undisturbed materials. The rate of the remolded verticalswell ratio is reflected in the heave prediction for the clayey
horizontal strips that were processed and compacted on site.

log (Po/98.07)=ao+aLLLL+aWW+ aD(D/9.81)(1)

In (1), ao, aLL, aW, and aD denote the regression coefficients offour independent variables. Their values are given in the row
labeled "[1]" in Table I. In this row, the coefficient ofdetermination (R2), taken from [1], is also presented. Table I
also contains the regression coefficients published in 1985 [2].
These coefficients were determined for the same 125 samples
as in [1] through a semi-statistical analysis, which assumed thatthe ratio aW/aD had a fixed value of 0.04. Obviously, no value
can be determined for R2 in such an analysis.
Table I: Regression coefficients of (1) and its R2
Regression coefficients of (1)
R2
Ref.
aW
aW
aLL
ao
[1]
0.66500.0208 *–2.1320 0.36
–0.0269--[2]
1.00000.0200–0.0400–2.0000[N]
0.65820.02020.38–0.0248–1.8823* The value of ao published in [2] as a quotation from [1] issurprisingly different: –1.868.
For the current study, a new multiple linear regression
analysis in the form of (1) was performed for the same 125
undisturbed samples. The results are listed in the row labeled"N" in Table I. This analysis yielded a standard error (SE) of
0.48. The significance of this value is that 68 percent of all257

predictions from (1), together with the data of row "N" of
Table 1, are expected to be accurate within a range of 10-0.48(i.e., 1/3) to 100.48 (i.e., 3) times the value calculated. Similarly,
95 percent of these predictions are expected to be accurate
within 0.1 up to 9 times the calculated value. Both these ranges
indicate the existence of an extensive dispersion characterizing
the measurement results, as do the low value obtained for R2. In
addition, it is important to emphasis that the results listed inrow "N" in Table I are not fully identical with those obtainedby [1]. Fig. I shows the gaps associated with these outputs.
300

LL=90%

250

Equality Line

6200

Reference [2]Equation (3)Equations (4) & (5)5150

W/PL=0.8

LL=70%

D=15 kN/m3

Vertical swell rate [%]Calculated Po according to [1] or [2] [kPa]

D=12, 14, 15, 16 kN/m3 and W/PL= 0.8, 1.0, 1.2

from different sites in Israel. The regressive equation
obtained is as follows:
Sp=–20.926+9.304log(LL)–1.948W/PL++4.250(D/9.81)–2.352log(Pp/98.07)(3)
2
For this regression, the R value obtained is 0.36, and the SE
value is 1.49%. In other words, 68 percent of all predictions
from (3) are expected to be accurate within ± 1.49%, and 95
percent of these predictions are expected to be accurate within± 2.98%. Fig. 2 depicts (3) graphically for the clay data
shown in the figure. For comparison purposes, this figure also
includes a graphical representation of (2) for the same clay.
Finally, the figure indicates that (3) leads to excessive values
for the swell pressure.

100
LL=50%
LL=50, 70, 90%
50
After [1]After [2]

0
0

50

100

150

200

250

3

LL=90%
2
70%

300

Calculated Po according to row labeled "N" in Table I [kPa]1

Figure 1. Display of the gaps in swell-pressure outputs
following (1) and the coefficients of Table I for the data shown
In Fig. 1, PL=20%, 25% and 30% for LL=50%, 70% and 90%,
respectively. For these data, the figure shows that the gap
between the current regression and the original regression
described in [1] is substantial. The ratio obtained for theoriginal regression outputs to the current regression outputs isabout 0.6. It is possible that this inexplicable gap may be dueto the existence of possible typing errors in the data publishedin [1]. As for the regression presented in [2], the gap obtained
is not constant, and the ratio varies from 0.6 up to 1.3.
In contrast to [1], the predictions of vertical swell rates (Sp)are given for the first time in [2]. For these predictions, Sp (inpercentages) depends on the applied vertical pressure, Pp (inkPa), and the defined characteristics of the undisturbed clay
samples. The formulation of the relevant equation is asfollows:
Sp=–3.683(Po/98.07)log(Pp/Po)

4(2)

In (2) the value of Po is determined from both (1) and the dataof the row labeled "[2]" in Table I. Here it is important to notethat the factor of -3.683 that appears in the equation isdetermined after the analysis of a series of empirical
relationships reported in the technical literature. The value of
this factor, unfortunately, did not receive any verificationfrom requested local laboratory tests, and hence this also is anindication of its weakness.
3. NEW GENERAL SWELLING MODELS
Reference [3] presents the new swelling model, based on amultiple linear regression performed on 352 new local
swelling test results obtained for undisturbed samples taken

50%
0
0

50

100

150

200

250

300

350

400

Vertical pressure [kPa]

Figure 2. Graphical comparison of three calculated variationsof ASTM vertical swell rates with vertical pressure for thedata shown
The same 352 undisturbed samples were also used todetermine the direct dependence of the variables in the samemanner described in (1) and (2). This was performed with theExcel-Solver command, leading to the following two
equations:log(Po/98.07)=–3.256+1.540log(LL)––0.537W/PL+0.738(D/9.81)(4)
Sp=–1.872(Po/98.07)log(Pp/Po)(5)
The SE value associated with the development of these two
equations is 1.45%, which is lower than that associated withthe development of (3): i.e., 1.49%. In addition to thisadvantage, another advantage of the model formulated in (4)and (5) is that they produce reasonable swell-pressure values
similar in principle to those obtained from the equations in [2].
For this issue, see also the numerical example in Fig. 3. All
these facts serve as a necessary justification for replacing the
LL independent variable by log(LL), and the W independent
variable by W/PL.
Furthermore, it is worthwhile examining the justification forusing (4) and (5), in which the LL independent variable has
been replaced by log(LL) and the W independent variable by
W/PL in the following way. First, there is need to formulate
two new equations with the aid of the Excel-Solver commandfor the 354 new undisturbed samples. This formulation,
which is obviously based on the same structure as (2) and (3),258

leads to the following two equations:log(Po/98.07)=–0.8936+0.0100LL––0.01576W+0.5251(D/9.81)(6)
Sp=–1.4836(Po/98.07)log(Pp/Po)(7)
It should be noted that (6) and (7) yield an SE value of 1.52%.
A graphical comparison of (4) and (5) (in which theindependents variables are, among others, log (LL) andW/PL) with (6) and (7) (in which the independents variables
are, among others, LL and W) is presented in Fig. 3 for thedata shown. This graphical representation also contains thetwo equations given by [2].
6Reference [2]Equations (4) & (5)Equations (6) & (7)5Vertical swell rate [%]W/PL=0.8
D=15 kN/m3

4

3

LL=90%

2
70%
1

50%

0
0

50

100

150
200
250
Vertical pressure [kPa]

300

350

400

Figure 3. Additional graphical comparison of three calculated
variations of ASTM vertical swell rates with vertical pressure
for the data shown
Figure 3 indicates that the use of the independent variables oflog (LL) and W/PL leads unexpectedly to a stronger effect of
the liquid limit on the vertical swell rate than that of the
independent variable of LL and W. In fact, this result is morenearly similar to the original equations presented by [2].
Finally, it is also important to note that the preference forreplacing W with W/PL arises from the fact that theacceptable characteristic of the moisture state that determines,
among other properties, the intensity level of the swelling
potential is W/PL [5].
In addition, the preference for the independent variables oflog (LL) and W/PL on LL and W, respectively, also derives
from the following comparison of the SE values: 1.45%
associated with (4) and (5), and 1.52% associated with (6) and(7). Finally, it can be concluded from the comparisonspresented in Fig. 2 and Fig. 3 that (4) and (5) are the mostappropriate to serve as the updated equations to define thegeneral swelling model in relation to the older equations
described in [2]; this is due to the fact that (4) and (5) yieldhigher results than do (6) and (7), thus giving the former pairof equations the benefit of the factor safety.
4. NEW CBR GENERAL SWELLING MODEL
Reference [4] describes several predicting models of thevertical swell rates measured in local CBR molds. The
samples of these molds are soaked in water and subjected toswell under a given surcharge pressure for a period of 4 days.

Obviously, these vertical swell rates are related to compacted(remolded) material. At this juncture it is important to note that
all the CBR swelling models of [4] do not belong to the same
formulation family of the new general swelling models as
expressed in (4) and (5). Thus, there is a need to formulize a
new CBR general swelling model with a similar formulation of
the above equations. Performing the Excel-Solver commandon the local 674 samples led to the following two equations:log(Po/98.07)=–0.8002+0.8625log(LL)––0.9370W/PL-0.0711(D/9.81)(8)
Sp=–6.0347(Po/98.07)log(Pp/Po)(9)where Sp denotes the CBR vertical swell, in percentages; Pp
denotes the vertical surcharge pressure, in kPa; LL denotes
the liquid limit, in percentages; PL denotes the plasticity limit,
in percentages; W denotes the water content, in percentages;
and D denotes the dry density, in kN/m3.
The SE value obtained for these two equations is 1.406%.
Here it should be noted that these two equations lead, contrary
to what was expected, to decreased values of Sp withincreasing values of D. This behavior stems from the way the
CBR test is conducted. As we know, the CBR molds are
soaked in water for four days, an operation that does not
ensure the complete saturation of the sample, unlike in thestandard ASTM 4546 testing.
In the CBR mold, the sample saturation occurs only at its top,
the depth of this saturated strip depending on the rate of thewater flow. When the dry density of the sample increases, therate of flow deceases, thus leading to decreasing thickness ofthe top of the swelling strip. Therefore, the combination of thisthickness reduction, which acts to decrease Sp, and theincreasing potential for swelling that occurs with increased dry
density creates the final balanced phenomenon in accordancewith the weighting influence of these two magnitudes. As
noted, this final phenomenon consists of a decline in the rate ofheave of the sample and, hence, in the rate of vertical swell. In
any case, the impact of D input on Sp output in (8) and (9) is
secondary as will be demonstrated later in this section.
To illustrate the issue of differences in the depth of thesaturation in the CBR molds after four days of soaking, Fig. 4
presents the relationship between the difference in moisture
content of the top 1 inch of the sample from that of the total
sample (both after four days of soaking in water) as seen withthe ratio of the molding water content to the plasticity limit of
the clay. This figure refers to swellings of clay samples takenfrom seven different locations in the area of Israel’s major
airport.
Fig. 4 shows that the difference in wetting after soaking the
CBR samples in water depends on the ratio of the molding
moisture content of the clay to its plasticity limit. With anincreasing moisture-content ratio, the rate of the difference inwetting indicates a decrease. In the same manner, Fig. 5
depicts the relationship between this difference in wetting andthe sample's dry density multiplied by the expression(1+PL/100). The figure refers to the same clay samples as inFig. 4. With an increasing rate of dry density, the rate of the259

difference in wetting is also shown to increase; however, thisoccurs only when the upper swelling thickness apparently
decreases.
14

Difference of moistures after soaking [%]Y-axis: moitsure content of the top one inch mimus the overall moisure content
LL=52%; PL=22%
LL=53%; PL=23%
LL=56%; PL=24%
LL=56%; PL=24%
LL=57%; PL=24%
LL=58%; PL=26%
LL=80%;PL=31%

12

10value, it is necessary once again to implement the
Excel-Solver command. This time, implementation should be
without the dry-density variable; such an implementation
leads to the following two equations:log(Po/98.07)=–1. 0099+0.93023log(LL)––0.9409W/PL(10)
Sp=–5.5283 (Po/98.07)log(Pp/Po)(11)
The value of SE obtained for these two equations, 1.409%, isvery similar to that obtained for the previous two equations.
Also, the difference in the predictions of the vertical swellingrate, which derived from (8) and (9) minus that derived from(10) and (11), is low. For the numerical examples of Fig. 6,
this difference is about 10% or less for the range of a 2%
vertical swell rate and higher.

8

6

4

2

0
0.5

0.6

0.7

0.8

0.9

1.0

1.1

1.2

1.3

5. COMPARISON OF CBR AND ASTM MODELS

1.4

Ratio of moisture content with plasticity limit

Figure 4. Difference in moisture content of the top 1 inch fromthat of the total sample (both after 4 days of soaking) with theratio of molding water content to the plasticity limit of the clay
14
LL=52%; PL=22%
LL=53%; PL=23%
LL=56%; PL=24%
LL=56%; PL=24%
LL=57%; PL=24%
LL=58%; PL=26%
LL=80%; PL=31%

12

10

8

6
Equation (10) & (11): CBR samples
Equation (4) & (5): ASTM undisturbed samples

6

Equations of [2]

5

2

0
17

18

19

20

21

22

23

Dry density [kN/m3] multiplied by the factor (1+PL/100)

Figure 5. Difference in moisture content of the top 1 inch fromthat of the total sample (both after 4 days of soaking), with dry
density multiplied by (1+PL/100)

4

3
70%
2
50%
1

0

20

Predicted vertical swell from (8) and (9) [%]W/PL=0.8
D=15 kN/m3

LL=90%

4

Vertical swell rate [%]Difference of moistures after soaking [%]Y-axis: moisture content of the top one inch minus the overall moisure content

Fig. 7 presents a graphical comparison of the swelling models
calculated from the outputs of the Excel-Solver command
performed on the test results of (a) 674 CBR samples as
expressed in (10) and (11) and (b) 352 undisturbed samples of
the ASTM testing as expressed in (4) and (5), the latter
serving as the local general swelling model. This comparison
includes the old general swelling model taken from [2]; i.e.,(1) and (2) together with the data of row "N" in Table I.

Full points refer to w/PL=0.8
Empty points refer to w/PL=1.0

180

D=1.5 t/m3
16

Equality line

Equality line +10%
14
12
10
8
6

LL=50%
LL=50%
LL=70%
LL=70%
LL=90%
LL=90%

4
2
0
0

2

4

6

8

10

12

14

16

18

20

Predicted vertical swell from (10) and (11) [%]Figure 6. Comparison of rate of vertical CBR swell according
to (8) and (9) with that according to (10) and (11) for the data
shown in the figure
Now, in order to neutralize the opposite effect of thedry-density input on the output of the vertical swell calculated

50

100

150
200
250
Vertical pressure [kPa]

300

350

400

Figure 7. Graphical comparison of CBR swelling modelobtained from (10) and (11) with that of ASTM obtained from
(4) and (5), including [2]The aforementioned comparison refers to a dry density of 15
kN/m3 and a ratio of moisture content to plasticity limit of 0.8.
Fig. 7 shows that the predicted swell-pressure rates of the
CBR remolded samples obtained from (8) and (9) are lower
than those of the undisturbed ASTM samples obtained from
(4) and (5). Here, it is important to note that the acceptance of
this finding should be of a limited nature, as the behavior of
the CBR swelling model beyond the surcharge pressure ratesof 10 up to 20 kPa stems from the fact of their being in theextrapolation zone of the experimental data. To recall,surcharge pressure rates of 10 up to 20 kPa were applied in
the CBR tests of [4].
Fig. 8 presents a variation of the ratio of measured vertical
CBR swell to the predicted vertical ASTM swell according to260

(4) and (5) with the variation in measured vertical CBR swell.
The figure contains ratio values for three suction levels forwhich the vertical swell rate was calculated using (4) and (5).
The logic of including two suction levels together with 0.0
kPa is the possibility that after four days of soaking in water,
the CBR sample will not reach a full saturated state, thus
leading to a residual suction value in the sample. To recall, aresidual suction value of 30 kPa exists in a clayey subgrade
when, covered by a standard asphalt pavement, it reaches anequilibrium state of excess moisture absorption [5], [6].
6.0Y-avis: vertical swell in CBR remolded samples
devided by vertical swell in ASTM undistubed samples

Ratio of remolded verical swel rate

5.0

y = 0.31x
R² = 0.47

y = 0.41x
R² = 0.47

4.0

y = 0.23x
R² = 0.49

3.0

2.0

1.0

Ratio of vertical swell of 1.0

0.0
-2

0
-1.0

2

4

6

8

10

12

14

16

Suction of 0.0 kg/sq.cm
Suction of 0.1 kg/sq.cm
Suction of 0.2 kg/sq.cm

that the standard error value of (12) is lower than thatobtained for (10) and (11); i.e., SE=1.41%. Can one thenconclude that the CBR swelling model expressed in (12) is
the most appropriate one for implementation?
In order to examine the likelihood of (12) and (13) comparedwith that of (10) and (11), Fig. 9 depicts a graphicalcomparison of CBR swelling models obtained from (12) and(13) with that obtained from (10) and (11) for the data shownin the figure. This figure clearly indicates a non-physicalcompatibility of (12) being able to serve as a basic functional
equation for developing a forecasting model for the verticalswell rates. The swell-pressure rates obtained from (12) in theextrapolation zone are completely unrealistic. Moreover, (12)does not allow for vertical swell rates to be negative (i.e., tobe changed from heave to deflection) for surcharge pressure
values greater than the swell-pressure values. Thus, theconclusion given in [4], that the impact of surcharge pressure
values on the rates of the vertical swell is only of minor
intensity, does not comply with reality. Obviously, thisconclusion stems from the use of an inappropriate basic
functional equation for the conditions of the given problem,
and therefore its validity is very questionable.
6-2.0

Equatins (10) & (11)Equation (12)Equation (13)Figure 8. Variation in the ratio of the measured vertical CBR
swell to the predicted vertical ASTM swell according to (4)and (5) with the variation in measured vertical CBR swellFig. 8 shows that the value of the slope of the line of variation
decreases with the increase in the suction level. This factindicates an increase in the frequency of cases in which theswell ratio (as defined previously) is greater than 1.0. For asuction value of 0.0 kPa, the majority of the measured points
is below the swell ratio of 1.0. Finally, the conclusion to be
derived from Fig. 8 is that the vertical swell rates of remolded
samples are higher than those of the undisturbed samples,
provided that the final suction level in these two kinds ofsamples is identical.
6. COMPARISON OF CBR SWELLING MODELSAs stated in Section 4, all the CBR swelling models of [4],which are based on the swelling results of 674 remolded
samples, do not belong to the same formulation family of new
general swelling models as expressed in (3) or (4) and (5).
Thus, it is interesting to compare these models with thatdeveloped in the present paper in Section 4. The two existingmodels of [4] are these:
Sp=47.2124×[(0.0423×LL–0.001662×Pp)/ln(D/9.81)]Q
Q=0.0078×W2–0.1617×W+2.6491(12)
Sp=72.0419+0.1216×LL–0.02822×Pp–8.9911×ln(D/9.81)–0.4665×W(13)
For the CBR swelling model expressed in (12), the value
obtained for R2 was 0.680, and that obtained for SE was
1.40%. In the same manner, the value obtained for R2 was
0.655 for the CBR swelling model expressed in (13), and1.45% for that obtained for SE. What is interesting to note isCalculated CBR Vertical swell rate [%]Vertical swell rate measured in a CBR mold [%]5

W/PL=0.8
D=15 kN/m3

LL=90%
4
70%
3

90%
50%

70%
2

1
50%

50%

90%
70%

0
0

50

100

150

200

250

300

350

400

Vertical pressure [kPa]

Figure 9. Graphical comparison of CBR swelling models
obtained from (12) and (13) with that obtained from (10) and(11) for the data shown in the figure
Fig. 9 also indicates the non-physical compatibility of (13) toserve as a basic functional equation for developing aforecasting model for vertical swell rates. The linear variationof the rate of vertical swell with the rate of the verticalsurcharge pressure as specified by (13) and shown in thatfigure has no reference in the technical literature, which
presents a wide range of laboratory test results for thispurpose. In addition, (12) and (13) indicate that increasing thedry density variable decreases the expected rate of the CBR
vertical swell. To recall, the same phenomenon ischaracterized by (8) and (9). However, for (13) these
deceased rates are not negligible, 0.6% or more for adry-density increase of 1.0 kN/m3. As for (12), this decreased
rate is, in contrast to (13), negligible, about 0.1% for the same
dry density increase of 1.0 kN/m3.
Together with the above-mentioned reservations concerning
the CBR swelling models of (12) and (13) in [4], it should be261

noted that the gaps between the families of the three CBR
swelling models shown in Fig. 9 are reduced for the zone ofactual vertical surcharge pressure applied in the CBR verticalswell measurements (i.e., 10 to 20 kPa). The conclusion is,again, that the choice of a basic functional equation to insert
into a statistical analysis of any measured data is critical,
especially for use in predicting values that are outside theactual measurement range; that is, in the range ofextrapolation. Therefore, the selection of a calculative modelfrom a number of calculative alternatives only on theprinciple of minimum standard error (SE) or maximum
coefficient of determination (R2) is seriously questionable.
7. SUMMARY AND CONCLUSION
A summary of all the developments described in thepreceding sections is shown in Table II for the ASTM
undisturbed samples and in Table III for the CBR remolded
samples. The data in these two tables are given for thefollowing independent variables: log(LL) and W/PL; i.e.,
instead of LL and W. When the principle of a minimum
standard error (SE) is applied to the 352 ASTM undisturbed
samples, the preferred solution is given by (4) and (5).
Table II: Summary of SE for the ASTM undisturbed samples
Section
No.
3
3

Eqns.
No.
3
4, 5

Samples'Number352
352

Analysis
MethodLinear Regression
Excel-Solver

SE
%
1.49
1.45

Another point to which attention need to be paid is thepreference for replacing the LL and W variables with log (LL)and W/PL, respectively, as discussed in Section 3. This section
shows that the equations corresponding to (4) and (5), namely(6) and (7), lead to an SE of 1.52%, compared with the 1.45%
listed in Table II (see also Section 3). In addition, as obtained
from Fig. 3, the effect of the variation in the LL independent
variable on Sp is more severe when the independent variables
are log (LL) and W/PL. This finding stems from thecomparison with the variation in the LL independent variablewith Sp as calculated from the two equations given in [2].
Finally it should be added that the analysis of the CBR samples
cannot be used to determine the ratio of the rate of remolded
vertical swell with the rate of undisturbed vertical swell,defined as the remolded vertical swell ratio. It is well known
that the final moisture state reached in a swell test conductedaccording to ASTM 4546 on undisturbed samples ischaracterized by the presence of zero suction, while that for aswell test conducted on remolded samples according to the
CBR procedure is characterized by the presence of a suction
value greater than zero. Because of these different suction
values, it is not possible to include the above two series ofexperimental results (i.e., the ASTM and the CBR results) into
one basket of results. Thus, in order to evaluate the rate of theremolded vertical swell ratio, it is necessary to conduct a seriesof tests according to ASTM 4546 on remolded samples fromthe same source of clay in parallel with the undisturbed
samples.

Table III: Summary of SE for the CBR remolded samples
Section
No.
---*
4

Eqns.
No.
---*
10, 11

Samples'Number674
674

Analysis
MethodLinear Regression
Excel-Solver

8. ACKNOWLEDGMENT

SE
%
1.63
1.41

* Taken from another local study [7].
Table III refers only to the equations that do not contain thedry density independent variable. This table indicates, again,that when the principle of a minimum standard error (SE) isapplied, the preferable solution is that of (10) and (11) as
derived from the Excel-Solver command. Again, for this case,the application of the above principle is legitimate, as thebasic functional equations are compatible with the physicalbehavior of the swelling phenomenon.
It is clear that the preference for (12) taken from [4], forwhich the SE value is 1.40% (i.e., less than the 1.41%
specified in Table III), is unacceptable, because the basic
functional equation of (12) that is used in the implementationof the statistical analysis of the reported measured data does
not represent the actual physical behavior of the variation in
the vertical swell rate with the vertical surcharge pressure rate.
As noted, the use of inappropriate basic functional equations
is particularly severe in the extrapolation zone of themeasured data. To conclude, the analysis of the measured
CBR vertical swell rates of the 674 remolded samples
presented in this paper can be used as an example ofpermissible and forbidden operations in order to determine
models for any measured data.

The statistical analysis of the data referred to in [4] was enabled
after obtaining all the raw data from Dr. Michael Divinsky,
former chief geotechnical statistician of the National Company
for Roads in Israel (PWD), to whom thanks are due.
9. REFERENCES[1] Komornik, A., and David, D., "Prediction of Swelling Pressure of Clays,"
Journal of Soil Mechanics and Foundation Division, Vol. 95, No. SM1,
1969, pp. 209-225.
[2] Wiseman, G., Komornik, A., and Greenstein, J., "Experience with Roads
and Buildings on Expansive Clays," Transportation Research Record
1032, 1985, pp. 60-67.
[3] Livneh, M., "Additional Updates in the General and Local Swelling
Model," Traffic and Transport, Journal of the Israel Section for Traffic
and Transportation, No. 98, March 2011, pp. 38-44 (in Hebrew).
[4] Ishai, I., Sahar, I., and Divinsky, M., "A Practical and Rational Model for
Swell Prediction of Clayey Subgrades Using the CBR Test," Paper
submitted to the 2004 Annual Meeting of the Transportation Research
Board, unpublished, 2004.
[5] Kassiff, G., Livneh, M., and Wiseman, G., Pavements on Expansive Clays,
Jerusalem Academic Press, 1969.
[6] Wiseman, G., Livneh, M., and Uzan, J., "Performance of a Full-Depth
Asphalt Pavement on Expansive Clay Subgrade," in Proc. of the 3rd Conf.
of the Road Engineering Association of Asia and Australia, Taipei, Taiwan,
1981, pp. 747-760.
[7] Livneh, M., "On the Updated Development of the General Swelling
Model," in Proc. of the 7th Conf. on Civil Engineering (CD Proc.), Israel
Association of Engineers for Construction & Infrastructures, Tel Aviv,
2011 (in Hebrew).

262

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Effect of Clay, Lime and Rice Husk Ash Contentson the Sorption Capacity of Cu, Ni and Zn by Soils
Pongsakorn Punrattanasin and Noppadol Sangiumsak
Research Center for Environmental and Hazardous Substance Management, Khon Kaen University andDepartment of Civil Engineering, Faculty of Engineering, Khon Kaen University, Thailand

ABSTRACT: Copper (Cu), nickel (Ni) and zinc (Zn) are
the common heavy metals that have been discharged fromindustries. The toxic conditions are always occurred withplants, animals, and human from these metals. This paper
presents the sorption of Cu, Ni and Zn by soils collectedfrom three different locations in the upper part ofNortheastern, Thailand. Then the new method was used toincrease the sorption capacity of Cu, Ni and Zn by the soils.
Clay, lime and rice husk ash were then selected and used as
the additional materials to improve the sorption capacity.
The soils were mixed with different ratios of the additional
materials. Batch equilibrium method was used as tools toinvestigate the adsorption value. The results showed thatthe adsorption fit to Langmuir Isotherm Model. All
additional materials can dramatically improve the sorption
capacity of Cu Ni and Zn by soils.

through sieve No.16. They can be classified by the Unified
Soil Classification System (USCS) as CL, SM and ML.
They will then be called CL-Soil, SM-Soil and ML-Soil,
respectively throughout this paper. Engineering properties
and porous properties of the soil samples are tabulated andshown in Table 1. The porous properties were evaluated byadsorption of nitrogen gas (N2) at -196 oC. The specificsurface area was calculated by Brunauer-Emmett-Teller(BET) equation. Figs.1 to 3 graphically show the SEM
photographs of CL-Soil, SM-Soil and ML-Soil,
respectively. As shown in Figs 1 to 3, it can be seen that thesurface of soil particles are almost rough.
Table 1 Engineering properties of soil

Keywords: Heavy metals, Adsorption, Isotherm
1. INTRODUCTION
Rapid growth of industries and communities in every partof Thailand including Northeastern leads to theenvironmental problem especially waste water andcontaminated soil. Many houses and industries discharged
the waste water into river or soil without treatment. This
resulted in an accumulated of toxins in soil and water.
Heavy metals are one kind of the toxins that has caused
serious health problem to humans and animals. They cause
damage to the kidney, liver, reproductive system [1],nervous system and ultimately death [2]. Ministry ofIndustry Thailand has enacted laws to limit the amount ofheavy metal in discharged waste water from industries.
There are many methods for remove the heavy metals fromthe waste water such as a reduction process, magnetic
ferrites treatment, ion exchange, reverse osmosis, andchemical oxidation and reduction, membrane separation,
coagulation, flotation, filtration, evaporation etc. [3,4,5].
Adsorption is one of the most common methods for heavy
metal removal because it is an effective and economical
technique. In this work, the adsorption of Cu, Ni and Zn bysoils found in the upper part of Northeastern, Thailand wasinvestigated and also the method of improvement of theadsorption by mixing soils with clay, lime and rice husk ash
was studied. The objective of this study is to find the effectof additional materials on the sorption capacity of soils.
2. MATERIALS AND METHODS
2.1 Adsorbent
The materials used as adsorbent were soils collected fromthree different locations in the upper part of Northeastern,
Thailand. They were excavated at different depths ranging
from 20-50 cm from the soil surface. The soils were dried at
a temperature of 110 oC for 48 hours and were sieved

Soil
Specific gravity
Optimum moisture
content (OMC), %Maximum dry unit
weight (  d ), kN/m3

CL-Soil
2.64

SM-Soil
2.60

ML-Soil
2.67

19.81

12.74

18.48

15.70

18.34

15.60

Coefficient of
permeability (k), cm/s
Specific surface area
(SBET), m2/g
Micropore volume (VDR),
cm3/g
Total pore volume (VT),cm3/g
Average pore diameter(DP), nm

0.59
x10-6

4.9
x10-6

8.6
x10-6

74

6

15

0.029

0.002

0.006

0.11

0.03

0.05

6

17

12

2.2 Additional Materials
The additional materials used to mix with soil for improve
the adsorption capacity in this study were rice husk ash,lime and clay. Rice husk ash and clay were taken from thelocal brick manufacturing plant and from Roi-Et GreenPower Plant, respectively. The lime or calcium oxide (CaO)
was bought from a construction material shop. All of theadditional material were dry at a temperature of 110 oC andwere then sieved through sieve No. 16. Figs.4 to 6 show theSEM photographs of rice husk ash, lime and clay,
respectively.
2.3 Heavy Metal Solutions
Heavy metals chosen as the adsorbed materials in this study
were Cu, Ni and Zn. They were prepared by dissolving
Copper Nitrate (Cu(No3)2), Nickel Nitrate (Ni(No3)2) andZinc Nitrate (Zn(No3)2) in distilled water. The solutions
were varied with concentrations of 5, 25, 50, 100, 250 and
500 mg/L.

263

(Fig.8). From the step above, it can be measured theadsorption equilibrium time and the amount of heavy metal
adsorbed by soil at equilibrium time (q). The equilibrium
amount of heavy metal can be calculated from thefollowing equation:q(Co  Ceq )Vsol

(1)

Ms

Where q is the adsorption capacity of the soil at equilibrium
condition (mg/g), Co is the initial concentration of heavy
metal solution (mg/g), Ceq is the equilibrium concentrationof the solution (mg/g), Vsol is the volume of solution (cm3),and W is the mass of soil (g). After the adsorption volume
was determined, the adsorption isotherm can then be
investigated. The adsorption isotherm is the relationshipbetween the concentration of the heavy metal solution atequilibrium condition (Ceq) and the equilibrium amount ofheavy metal adsorbed by soil (q). In this study, theLangmuir Isotherm Model was used to describe theadsorption isotherm which represented by the following
equation:Fig.1 SEM photograph of CL-Soilq

 Ceq1  C eq(2)

Where  is Langmuir constant related to the energy (L/mg),
 is the adsorption capacity (mg/g).

Fig.2 SEM photograph of SM-Soil

Fig.4 SEM photograph of rice husk ash
Fig.3 SEM photograph of ML-Soil

2.4 Experimental work
In this study, batch adsorption test was carried out toinvestigate the heavy metal adsorption. The test processes
can be performed by mixing a 2.5 g of soil with the Cu, Ni
and Zn solutions. The initial concentrations were ranged
from 5 to 500 mg/L and put it in 120 mL plastic bottle. Thenthe mixture was shaken by the horizontal shaker with aspeed of 130 cycles per minute for 0.5, 1, 3, 6, 12, 24, 48,
and 72 hours as shown in Fig.7. After a particular period of
the time, the soil was separated from the heavy metal
solution by using a 0.45µm filter and diluted it into thesolution. Finally, concentration of the residual solution was
analyzed by using Atomic Adsorption Spectrometer (AAS)
264

Fig.5 SEM photograph of lime

results was also found when performing the tests with Cu,
Ni and Zn for all soils.
100

CL-Soil
SM-Soil80

Ct (mg/L)

ML-Soil60

40

20

Fig.6 SEM photograph of clay
0
0

12

24

36

48

60

72

Time (hr)Fig.9 Effect of contact time on Cu adsorption by CL-Soil,
SM-Soil and ML-Soil
100

CL-Soil
SM-Soil80

Ct (mg/L)

ML-Soil60

40

20

Fig.7 Shaking the mixture of soil and heavy metal solutions

0
0

12

24

36

48

60

72

Time (hr)Fig.10 Effect of contact time on Ni adsorption by CL-Soil,
SM-Soil and ML-Soil
100

CL-Soil80

SM-SoilCt (mg/L)

ML-Soil

Fig.8 Atomic Adsorption Spectrometer (AAS)

60

40

20

3. RESULT AND DISCUSSION
3.1 Equilibrium time of adsorption
The adsorptions of Cu, Ni and Zn for an initialconcentration of 100 mg/L by CL-Soil, SM-Soil and
ML-Soil at different time were shown in Figs.9 to 11. The
results show the remaining concentration of the heavy
metal (Ct) as a function of time. It can be seen that theadsorption took place rapidly at the beginning of thereaction, which the concentration rapidly decreased at theperiod of 1-3 hours and decline over time until reaching theequilibrium condition within 6-12 hours. The same trend of

0
0

12

24

36

48

60

Time (hr)Fig.11 Effect of contact time on Zn adsorption by CL-Soil,
SM-Soil and ML-Soil
3.2 Adsorption Isotherm
Figs.12 to 14 showed the adsorption isotherm of Cu, Ni and
Zn for different soil samples (CL-Soil, SM-Soil and
ML-Soil). As shown in Figs.12 to 14, the amount ofadsorption (q) increased with increasing equilibrium

265

72

concentration (Ceq). The adsorption isotherm can be
represented by Langmuir adsorption isotherm except theadsorption of Cu by ML-Soil (as showed in Fig. 12). The
consistency of the Langmuir isotherm describes that thesurface of soils was covered with monolayer of the metal
particles [5]. Langmuir parameters ( and ) and thecoefficient of correlation (R2) were then summarized inTable 2. As shown in Table 2, SM-Soil has the lowest
adsorption capacity () when performing the tests with Cu,
Ni and Zn compared to CL-Soil and ML-Soil. This can be
explained from the results that SM-Soil has the lowest
specific surface and the lowest surface roughness than
CL-Soil and ML-Soil (data shown in Table 2 and Figs.1-3).
Very high coefficient of correlation (R2) indicated that theadsorption of Cu, Ni and Zn by all soil samples favorable fit
to Langmuir Isotherm. When comparing the adsorption ofeach heavy metal, it can be seen that the adsorption capacity
was in the order of Cu>Ni>Zn.

Table 2 Langmuir parameters of soil samples
Langmuir Parameters
Solution
Soil
R2
 (mg/g)  (L/mg)
CL-Soil
4.88760.1466
0.9858Cu
SM-Soil
2.93430.1161
0.9948
ML-Soil
1.17231.32250.8507
CL-Soil
3.88050.03710.9310Ni
SM-Soil
1.84230.02960.9903
ML-Soil
4.30480.01980.9354
CL-Soil
3.02850.03320.9855Zn
SM-Soil
1.22400.06100.9985
ML-Soil
2.63230.16840.9992

10

CL-Soil
SM-Soil

8

ML-Soilq (mg/g)

10

CL-Soil
SM-Soil864q (mg/g)

ML-Soil
6

2

4

0
0

100

200

300

400

500

Ceq (mg/L)
2

0
0

100

200

300

400

500

Fig.14 Adsorption isotherm of Zn adsorption by CL-Soil,
SM-Soil and ML-Soil

Ceq (mg/L)

Fig.12 Adsorption isotherm of Cu adsorption by CL-Soil,
SM-Soil and ML-Soil
10

CL-Soil
SM-Soil8q (mg/g)

ML-Soil
6

4

2

0
0

100

200

300

400

500

Ceq (mg/L)

Fig.13 Adsorption isotherm of Ni adsorption by CL-Soil,
SM-Soil and ML-Soil

3.3 Effect of the additional material on the adsorptionof heavy metals
Because the SM-Soil has the lowest adsorption of heavy
metal (as already shown in details in the previous section),
this soil was chosen as a representative of all soils to findthe method of improvement of the adsorption capacity. The
SM-soil was replaced with the rice husk ash, lime and clayat the amount of 0, 20, 40, 60, 80 and 100% by weight. The
effects of the three additional materials on the adsorption
capacity of Cu, Ni and Zn are presented in Figs.15 to 17,
respectively. As presented in Figs.15 to 17,
all theadditional materials especially lime can dramatically
improve the adsorption capacity of heavy metals (q). The
adsorption capacity increased with increasing amount ofrice husk ash and clay content. While using 20% of lime in
the replacement, the results showed the maximum
adsorption capacity of Cu and Ni. In case of Zn, theadsorption almost reached the maximum when 40% of lime
was used in the replacement. It was possible to conclude
that the proper amount of lime which should be used toreplace SM-Soil was 20% for increasing the adsorption
capacity of Cu and Ni and 40% for adsorption capacity of
Zn. The optimum content and the recommended amount ofrice husk ash, lime and clay using to mix with SM-Soil aresummarized in Table 3. The recommended contents may
not be the optimum contents. The optimum contents were
the amount of the additional material that provided the266

maximum adsorption capacity. The recommended contents
were evaluated by the consideration both the volume of
replacement and the adsorption capacity.

43

In this study, the adsorption capacities of Cu, Ni and Zn bythree kinds of soil (CL, SM and ML) collected from threedifferent locations in the upper part of Northeastern,
Thailand were investigated. Further, the new method wasintroduced to improve the adsorption capacity of soil bymixing the soil with rice husk ash, lime and clay. The result
indicated that the adsorption reaches the equilibrium timewithin 6-12 hours. The adsorption capacity increased withincreasing initial concentration. Langmuir isotherm model
can favorable describe the adsorption. The adsorption
capacity of all soil was in the order of Cu>Ni>Zn. The
SM-Soil provided the lowest adsorption capacity because
the specific surface area and roughness were the importantfactors which affected on the adsorption. From theimprovement of the adsorption capacity, rice husk ash, lime,
and clay can increase the adsorption capacity of the
SM-Soil. Lime showed the highest performance for
improvement of Cu and Zn adsorption while the highest
performance for improvement of Ni was found by rice husk
ash. The recommended contents of lime was 20% by
weight of soil in the replacement for adsorption of Cu andNi and 40% by weight of soil in the replacement for
adsorption of Zn. For the rice husk ash, the recommended
content was 60% by weight of soil for adsorbing the Ni.

q (mg/g)

CONCLUSION

4

2

1

Rice husk ash
0

Clay

20

40

60

80

100

% Replacement

Fig.17 Effect of the additional material on Zn adsorption by
SM-SoilTable 3 The optimum contents and the recommended
amounts of rice husk ash, lime, and clay using to mix with
SM-Soil
Optimum Increasing Recommended
Additional Heavy
contentof q
Amountmaterials metals
(%)
(%)
(%)
RHA

Lime

4

Clay

3

q (mg/g)

Lime

0

Cu

100

225

-

Ni

100

223

60

Zn

100

183

-

Cu

20

253

20

Ni

20

141

20

Zn

100

210

40

Cu

100

126

-

Ni

100

125

-

Zn

100

114

-

2

5 ACKNOWLEDGMENT
1

Rice husk ash

Lime

Clay

0
0

20

40

60

80

100

% Replacement
Fig.15 Effect of the additional material on Cu adsorption by
SM-SoilThis research was funded by Research Center for
Environmental and Hazardous Substance Management,
Khon Kaen University. The authors wish to acknowledge
and grateful thank to Dr. Arthit Neramittagapong, Dr.
Yuvarat Ngernyen and Miss Wimonporn Iamamornphonthof the Department of Chemical Engineering, Khon Kaen
University.
6

REFERENCES[1]

Mohumed EI Zayat and Edward Smith, “Removal of heavy metals
by using activated carbon produced from cotton stalks,” Canadian
Journal on Environmental, Construction and Civil Engineering, vol.
1, September. 2010, pp. 71-79.
Y. B. Onundi, A. A. Mamun, M. F. Al Khatib and Y. M. Ahmed ,“Adsorption of copper, nickel and lead ions from synthetic
semiconductor industrial wastewater by palm shell activated carbon,”International Journal of Environmental Science and Technology,
vol. 7(4), 2010, pp. 751-758.
Souag R, Touaibia D, Benayada B and Boucenna A, “Adsorption ofheavy metals (Cd, Zn and Pb) from water using keratin powder
prepared from Algerien sheep hoofs,” European Journal of
Scientific Research, vol. 35, No.3. 2009, pp. 416-425.
Kailas L. Wasewar, “Adsorption of Metals onto Tea Factory Waste:
A Review,”IJRRAS, vol. 3(3), June 2010, pp. 303-322.
Tae-Young Kim, Sun-Kyu Park, Sung-Yong Cho, Hwan-Beom Kim,
Yong Kang, Sang-Done Kim and Seung-Jai Kim, “Adsorption ofheavy metals by brewery biomass,” Korean Journal, Chemical
Engineerin, vol. 22(1). 2005, pp. 91-98.

4[2]

q (mg/g)

3

2

[3]
1

Rice husk ash

Lime

Clay

[4]

0
0

20

40

60

80

100

[5]

% Replacement

Fig.16 Effect of the additional material on Ni adsorption by
SM-Soil0267

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Effect of Temperature on the Sorption Capacity of Copper by Soils
Noppadol Sangiumsak1 and Pongsakorn Punrattanasin2
Department of Civil Engineering, Faculty of Engineering, Khon Kaen University, Thailand
E-mail: g_noppadol@hotmail.com1, ponpun@kku.ac.th2


ABSTRACT: Copper is a common heavy metals which isdischarged from various industries. In previous studies, it canbe seen that soil is a potential adsorbent to remove heavy
metal from waste water. This paper presents the effect oftemperature on the sorption capacity of Copper by soils foundin the upper part of Northeastern, Thailand. The experiments
were done by using batch equilibrium test method. The
adsorptions were tested and measured at temperatures
between 30 to 50 degrees of Celsius and the initialconcentrations of heavy metals are in the range of 100 to 1000
mg/L. The results from the experimental study showed thatthe adsorption isotherms of the soils can be satisfactorily
described by Langmuir model. It is also found thattemperature plays an importance role to the sorption capacityof the heavy metals by soils. Higher temperature of heavy
metal solutions can make the higher sorption capacity.

from 20-50 cm from soil surface. The soils were dried in theoven at a temperature of 110 oC for 48 hours and then weresieved through sieve No.16. The results from sieve analysis
and Atterberg limits test demonstrated that the soil samples
can be classified as CL, ML, SM and SM-SC by the Unified
Soil Classification System (USCS). They will be named
according to their kind throughout this paper as CL-Soil,
ML-Soil, SM-Soil and SM-SC-Soil, respectively.
Engineering properties and porous properties of the soil
samples are tabulated and shown in Table 1. The porous
properties were evaluated by adsorption of nitrogen gas (N2)
at -196 oC. The specific surface area was calculated by
Brunauer-Emmett-Teller (BET) equation. Figs.1 to 4graphically show the SEM photographs of CL-Soil, ML-Soil,
SM-Soil and SM-SC-Soil, respectively. As shown in Figs 1 to4, it can be seen that the surface of all soil particles are almost
rough.

Keyword: Adsorption, Heavy metal, Soil, Temperature

Table1 Characteristics of soil samples
Characteristics

1. INTRODUCTION
Rapid growth of industries and municipalities in Thailand has
led to increase the environmental problem. A large amount ofwaste water has been discharged from houses and industries
into river or soil without treatment. The waste water often
contaminated with toxins. One kind of the toxins is heavy
metals. Various industries such as mining, plating, dyeing,
and metal processing released the waste water contaminatedwith heavy metal [1]. The heavy metals may be found in soils,
water, sediments and plants [2]. The heavy metals cause
damage to the kidney, liver, reproductive system [3] andnervous system [4]. There are many methods for remove theheavy metals from the waste water such as a reduction
process, magnetic ferrites treatment, ion exchange, reverse
osmosis, chemical oxidation, membrane separation,
coagulation, flotation, filtration, evaporation etc. [5,6,7].
Adsorption is one of the most common methods for heavy
metal removal because it is an effective and economical
technique. Many studies demonstrated that soil is a potential
adsorbent to remove heavy metal from waste water[8,9,10,11].
The objective of this study is to investigate the effects oftemperature on the adsorption capacity of Cu by soil found in
the upper part of Northeastern, Thailand. The adsorptions
were done at temperatures ranging from 30 oC to 50 oC.

Specific gravity
Coefficient of
permeability (k),cm/sSpecific surface area
(SBET), m2/g
Micropore volume(VDR), cm3/g
Total pore volume(VT), cm3/g
Average pore
diameter (DP), nm

2.64

2.67

2.66

SM-SCSoil
2.62

0.59
x10-6

8.6
x10-6

9.6
x10-6

14.9
x10-6

74

15

103

54

0.029

0.006

0.040

0.021

0.11

0.05

0.21

0.11

6

12

8

8

CL-Soil ML-Soil SM-Soil

2. MATERIALS AND METHODS
2.1 Adsorbent
Four soil samples collected from different locations in theupper part of Northeastern, Thailand were used as theadsorbent. They were excavated at different depths ranging

269

Fig.1 SEM photograph of CL-Soilstudy. The solution of Cu was prepared by dissolving copper
nitrates (Cu(NO3)2) in distilled water. The concentration of
Cu solution was varied with concentrations of 100, 250, 500
and 1000 mg/L. The properties of copper nitrates used aresummarized in Table 2.
Table 2 Properties of copper nitrate
Properties
FormulaMolecular weight (g/mol)
Density (g/cm3)
Solubility (g/100ml)

Copper NitrateCu(NO3)2
295.652.07
137.8

2.3 Experimental work
In this study, the adsorption of Cu with respect to contact time,
initial concentration and temperature was experimentallyinvestigated. The test process can be done by mixing a 2.5 gof soil with Cu solution. The initial concentrations are ranged
between 100 to 1000 mg/L. The temperature of the mixtures
was controlled at 30 oC , 40 oC and 50 oC by putting in thewater bath as shown in Fig. 5. The contact times were from 1
to 48 hours. After a particular period of the time, the soils
were separated from the Cu solution by using a 0.45µm filter
and the solution were then diluted to limit a concentration notexceed 5 mg/L. Finally, concentration of the residual solution
was analyzed by using Atomic Adsorption Spectrometer(AAS) (Fig.6). From the step above, it can be measured theadsorption equilibrium time and the amount of heavy metal
adsorbed at equilibrium time. The amount of heavy metal
adsorbed can be calculated from the following equation:

Fig.2 SEM photograph of ML-Soilq
Fig.3 SEM photograph of SM-Soil(C o  C eq )Vsol

(1)

Ms

Where q is the adsorption capacity of the soil at equilibrium
condition (mg/g), Co is the initial concentration of heavy
metal solution (mg/g), Ceq is the equilibrium concentration of
the solution (mg/g), Vsol is the volume of solution (cm3), andM is the mass of soil (g).

Fig.4 SEM photograph SM-SC-Soil
2.2 Heavy Metal Solutions
Cu is one kind of the most heavy metals, which generally
found in the discharged wastewater from industries andhouses. It was then chosen as the adsorbed materials in this270

Fig.5 Temperature control of the adsorption

500
30 C

40 C

50 C

ct (mg/L)

400
300
200
100
0
0

12

24
Time (hr)36

48

Fig.7 Adsorption of Cu by CL-Soil at different temperatures
500
30 C

Fig.6 Atomic Adsorption Spectrometer (AAS)

40 C

50 C

After the adsorption volume was determined, adsorption
isotherm was investigated. The adsorption isotherm is therelationship between the concentration of the heavy metal
solution at equilibrium condition (Ceq) and the equilibrium
amount of heavy metal adsorbed by soil (q). In this study, theLangmuir isotherm model was used to describe theadsorption isotherm which represented by the following
equation:q Ceq1  Ceq

ct (mg/L)

400
300
200
100
0
0

12

24
Time (hr)36

48

Fig.8 Adsorption of Cu by ML-Soil at different temperatures

(2)

500

Where  is Langmuir constant related to the energy (L/mg), 
is the adsorption capacity (mg/g)
ct (mg/L)

3.1 Equilibrium time of adsorption
The adsorptions of Cu at different times and different
temperatures for an initial concentration of 500 mg/L by
CL-Soil, ML-Soil, SM-Soil and SM-SC-Soil were shown in
Figs.7 to 10, respectively. The Figures show the remaining
concentration of the heavy metal (Ct) as a function of time. It
can be seen that the adsorption occurred rapidly at the time of
1 to 3 hours. The adsorption rate gradually decreased andeventually reached the equilibrium. It can be said that theadsorptions reached equilibrium within 6 hours for CL-Soiland SM-SC-Soil and within 24 hours for SM-Soil. For
ML-Soil, it seems that the adsorption still increased after 48
hours. It was found that the temperature greatly affected on
the adsorption of Cu by all soil. The adsorption increased withincreasing temperature. However, it can be observed that theequilibrium times did not depend on temperatures.

40 C

50 C

300
200
100
0
0

12

24
Time (hr)36

48

Fig.9 Adsorption of Cu by SM-Soil at different temperatures
500
30 C

40 C

50 C

400

ct (mg/L)

3. RESULT AND DISCUSSION

30 C
400

300
200
100
0
0

12

24
Time (hr)36

48

Fig.10 Adsorption of Cu by SM-SC-Soil at different
temperatures271

3.2 Adsorption isotherm
Figs.11 to 14 show the adsorption isotherm of Cu for CL-Soil,
ML-Soil, SM-Soil and SM-SC-Soil, respectively. As shownin Figs.11 to 14, the adsorption capacity (q) increased withincreasing equilibrium concentration (Ceq). For CL-Soil,
ML-Soil and SM-SC-Soil, the increasing rate of q gradually
decreased and converged to the maximum value. In next step,
the relationships of Ceq versus Ceq/q were plotted as shown in
Figs.15 to 18 to determine the values of  and . The valuesof  and  can be determined from the slope and interception( =1 /slope,   1 / c where c = interception). It can befound that the adsorption isotherm of CL-Soil, ML-Soil and
SM-SC-Soil soils can be represented by Langmuir
Adsorption Isotherm. The consistency of the Langmuir
Isotherm describes that the surface of soils was covered withmonolayer of the metal particles [1]. SM-Soil show different
character of the adsorption isotherm. The increasing rate of qwith respect to Ceq of SM-Soil negligible decreased so that
Freundlich Isotherm is more proper than Langmuir Isotherm
in describing the adsorption of SM-Soil. Langmuir
parameters ( and ) and the coefficient of correlation (R2)were summarized in Table 3. The last column of Table 3
shows the increasing value of  when the temperature
increased from 30oC to 40oC and 50oC. The improvement canbe calculated from the following equation:% Increasing of  =( T  30 )30

8

q (mg/g)

6

4

2
30 C
0

q (mg/g)

15
10

30 C

30 C

40 C

50 C

0

200

400
600
Ceq (mg/L)

800

1000

Fig.11 Adsorption isotherm of Cu by CL-Soilat different temperaturesq (mg/g)

40 C

50 C

0
200

400
600
Ceq (mg/L)

800

1000

Fig.13 Adsorption isotherm of Cu by SM-Soilat different temperatures5

q (mg/g)

4
3
2
1
30 C

40 C

50 C

0
0

200

400
600
Ceq (mg/L)

800

1000

Fig.14 Adsorption isotherm of Cu by SM-SC-Soilat different temperatures
4

0

1000

5

5

1

800

20

temperature of 30 oC, respectively. The temperature of 30 oC
is the representative normal temperature in the Northeastern,
Thailand. As shown in Table 3, the adsorption capacity ()significantly increased with increasing temperature. The
effect of temperature occurred with ML-Soil and SM-SC-Soil.
It can be found that the increasing of  can reach up to 25.6%
when the temperature increased from 30oC to 50oC. The
results of adsorption of CL-Soil, ML-Soil and SM-SC-Soilshow a very good result from the high value of R2. This
indicated that the adsorption of Cu by these soils well fit toLangmuir Isotherm Model. Although the adsorption isotherm
is not fit to Langmuir Isotherm, it can see from Figs. 9 and 13
that the adsorption of Cu by SM-Soil was still very high.

2

400
600
Ceq (mg/L)

25

Where T and  30 are the  values at temperature T and at

3200

Fig.12 Adsorption isotherm of Cu by ML-Soilat different temperatures

0

4

50 C

0

(3)

x100

40 C

CONCLUSION

In this study, the effects of temperature on the adsorption
capacity of Cu by four different kinds of soils (CL, ML, SM
and SM-SC) found in the upper part of Northeastern,
Thailand were investigated. The experiments were done byusing batch equilibrium method. The temperature was varied
over the range of 30 to 50 oC. The results revealed that theadsorption occurred very rapid at the period of 1 to 3 hours
and reach to the equilibrium condition with different times.
The equilibrium condition occurred within 6 hours for
CL-Soil and SM-SC-Soil and occurred within 24 hours for
SM-Soil. The temperature do not affect to the equilibrium272

time. The adsorption of CL-Soil, ML-Soil and SM-SC-Soil fit
to Langmuir Adsorption Isotherm very well. The adsorption
capacity significantly increased with increasing temperature.
For the ML-Soil and SM-SC-Soil, the adsorption capacity of
Cu increased up to 25.6% when the temperature increased
from 30oC to 50oC.

300
250

Ceq/q (mg/g

200
150
100

0

200

Ceq/q (mg/g

30 C
40 C
50 C

50

250

0

200

150

800

1000

Fig.18 Langmuir Adsorption Isotherm of ML-Soilat different temperatures

100
30 C
40 C
50 C

50

Table 3 Summarized of Langmuir parameters of soil samples
Langmuir Parameters

0
0

200

400
600
Ceq (mg/L

800

Soil

1000

140
120
100
80
60
40

30 C
40 C
50 C

20
0
0

200

400
600
Ceq (mg/L

800

1000

Fig.16 Langmuir Adsorption Isotherm of ML-Soilat different temperatures

Temp.

30 oC
CL
40 oC
50 oC
30 oC
ML
40 oC
50 oC
30 oC
SM
40 oC
50 oC
30 oC
SM-SC 40 oC
50 oC

Fig.15 Langmuir Adsorption Isotherm of CL-Soilat different temperatures

Ceq/q (mg/g

400
600
Ceq (mg/L

5

 (mg/g)  (L/mg)
3.81
4.12
4.52
5.76
6.40
7.24
17.70
32.36
144.93
3.16
3.53
3.97

0.09
0.07
0.05
0.16
0.18
0.15
0.04
0.05
0.01
0.10
0.06
0.04

R2
1.00000.99990.99930.99991.00001.00000.96470.91140.09180.99990.99900.9982%
Increasing
of 
0
8.2
18.8
0
11.0
25.6
0
0
11.7
25.6

ACKNOWLEDGMENT

This research was funded by Faculty of Engineering, Khon
Kaen University. The authors wish to acknowledge andgrateful thank to Dr. Arthit Neramittagapong, Dr. Yuvarat
Ngernyen and Miss Wimonporn Iamamornphonth of the
Department of Chemical Engineering, Khon Kaen
University.

Ceq/q (mg/g

14
12

6

REFERENCES

10

[1]

Tae-Young Kim, Sun-Kyu Park, Sung-Yong Cho, Hwan-Beom Kim,
Yong Kang, Sang-Done Kim and Seung-Jai Kim, “Adsorption of heavy
metals by brewery biomass,” Korean Journal, Chemical
Engineering, vol. 22(1). 2005, pp. 91-98.
A. Dube, R. Zbytniewski, T. Kowalkowski, E. Cukrowska, B.
Buszewski, “Adsorption and Migration of Heavy Metals in Soil,”Polish
Journal of Environmental Studies, vol. 10, No. 1, August 2000, pp.
1-10.
Mohumed EI Zayat and Edward Smith, “Removal of heavy metals byusing activated carbon produced from cotton stalks,” Canadian
Journal on Environmental, Construction and Civil Engineering, vol. 1,
September. 2010, pp. 71-79.
Y. B. Onundi, A. A. Mamun, M. F. Al Khatib and Y. M. Ahmed ,“Adsorption of copper, nickel and lead ions from synthetic
semiconductor industrial wastewater by palm shell activated carbon,”International Journal of Environmental Science and Technology, vol.
7(4), 2010, pp. 751-758.
Souag R, Touaibia D, Benayada B and Boucenna A, “Adsorption ofheavy metals (Cd, Zn and Pb) from water using keratin powder

8
6

[2]

4

30 C
40 C
50 C

2

[3]

0
0

50

100
Ceq (mg/L

150

200
[4]

Fig.17 Langmuir Adsorption Isotherm of ML-Soilat different temperatures
[5]

273

prepared from Algerien sheep hoofs,” European Journal of Scientific
Research, vol. 35, No.3. 2009, pp. 416-425.
[6] Kailas L. Wasewar, “Adsorption of Metals onto Tea Factory Waste: A
Review,”IJRRAS, vol. 3(3), June 2010, pp. 303-322.
[7] Jihyun Lim, Hee-Man Kang, Lee-Hyung Kim and Seok-Oh Ko,“Removal of Heavy Metals by Sawdust Adsorption: Equilibrium andKinetic Studies,”Environmental Engineering Research, vol. 13, No.2,
March 2008, pp. 79-84.
[8] Saad A. Al-Jlil, “Removal of Heavy Metals from Industrail Wastewaterby Adsorption using Local Bentonite Clay and Roasted Date Pits in
Saudi Arabia, ” Trends in Applied Sciences Research, vol. 5(2), 2010,
pp. 138-145.
[9] Islem Chaari, Mounir Medhioub and Fakher Jamoussi, “Use of Clay to
Remove Heavy Metals from Jebel Chakir Landfill Leachate,”Journal ofApplied Science in Environmental Sanitation, vol. 6, No.2, June 2011,
pp. 143-148.
[10] Chuangcham U and Charusiri P, “Adsorption of Heavy Metals from
Landfill Leachate in Soil: A Case Study of Kham Bon Landfill, Khon
Kaen Province, NE Thailand,”Proceedings of The International
Symposia on Geosciences and Environments of Asian
Terranes(GREAT 2008), 4th IGCP516 and 5th APSEG,2008, pp.
501-505.
[11] M. Ali Awan, Ishtiaq A. Qazi and Imran Khalid, “Removal of Heavy
Metals through Adsorption Using Sand,”Journal of Environmental
Sciences, vol. 15, No.3, 2003, pp. 413-416.

274

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Experimental Investigation for Nitrate Reduction Using Iron Powder in PorousMedia
Shimada H, Inoue K, Ihara I, Suzuki K and Tanaka T, Graduate School of Agricultural Science, Kobe University

ABSTRACT: This paper assessed the potential of iron powder as anitrate reduction and retardation material in porous media. Under
saturated and unsaturated flow conditions, column experiments wereconducted in silica sand with a mixture of two kinds of iron powders.
Experimental breakthrough curves were analyzed using temporal
moments not only to estimate the retardation factor and thedegradation rate but to quantify mass recovery fractions associatedwith nitrate and ammonium ions. The results showed the increase of
the amount of iron powder resulted in the increase of the degradation
rates and the retardation factor estimates. Moreover, the productionof ammonium occurred by nitrate reduction was confirmed. A
mixture of iron powder with soil has a potential to reduce theconcentration of nitrate by the reduction of nitrate to ammonium andto retard a nitrate.

Keywords: Nitrate, Reduction, Iron Powder, Column
Experiments
1. INTRODUCTION
Excessive nitrate concentrations in groundwater are related toland use activities including pastoral farming market
gardening, application and leaching of nitrogenous fertilizers
and industrial and sewage waste disposal [1]. Overloading
agricultural soils with nitrogen leads to rising nitrateconcentration in groundwater, while enhancing crop growthby the application of nitrogen gives economic benefits inmany parts of the world. Protection of groundwater from
nitrate contamination is an often overlooked health concern.
The immediate health concern is the reduction of nitrate tonitrite in the digestive tract by nitrate-reducing bacteria.
Nitrite is readily absorbed into the blood where it combineswith the hemoglobin that carries oxygen, leading to the formof methemoglobin, which is unable to transport oxygen, andthe physical stress. When severe enough, nitrate poisoning islife threatening because of suffocation, which is called
methemoglobinemia, or blue baby syndrome, in infants[2],[3]. Nitrate released to the environment can also increase
eutrophication levels in rivers or lakes.
In general, nitrate can readily be transported beneath the soil
zone without ion exchange or adsorption to soil surface.
Therefore, high concentration of nitrates can be found inshallow groundwater systems, especially in sandy aquifers[4],[5]. There has been developed many processes forremoving nitrate from water, including biologicaldenitrification, separation with ion exchange or reverse
osmosis and reduction by chemical reductants [6]. However,almost all treatment processes are performed aftergroundwater extraction by wells or drains, which is theprimary technology to treat contaminated groundwater [7],and have a markedly different property from the reduction of

nitrate concentrations at a source in a contaminated site. In
addition, at a vast contaminated site groundwater extraction
might be ineffective [8]. This ineffectiveness has been led tothe development of alternative and unique approaches
towards groundwater cleanup [9]-[11].
At present, reduction of nitrate using zero-valent iron (Fe0 or
ZVI) is a highly exergonic reaction and is expected to be analternative for nitrate removal from water [6]. ZVI has beenused to remediate contaminated groundwaters by placing
large quantities of granular metal to act as a permeable barrier[12]. Although nitrate reduction using ZVI is a rapid reaction
if the solution pH remains within an acidic range [13], theformation of passivating scales on the ZVI over time may
limit its long-term reduction potential since nitrate reduction
by ZVI is a corrosive process. Previous work has
demonstrated the ability for ZVI to reduce nitrate in batch
experiments, with nearly stoichiometric recovery of nitrogenas ammonium [14],[15]. However, only a few studies havebeen reported the reduction of nitrate and in column
experiments [16].
The present work is aiming to evaluate the reduction of nitrateand nitrate transformation to ammonium ions by ZVI incolumns where a sandy soil is homogeneously mixed withZVI and to quantify the retardation of nitrate onto soil and/or
ZVI surfaces. Temporal moment approaches are applied toestimate the retardation factor and the degradation constantbased on the experimental breakthrough curves (BTCs) incolumns. Specifically, the effect of the weight ratio of ZVI tosandy soils on the degree of the retardation and degradation of
nitrate is of concern to explore the possible use of ZVI in afield.
2. MATERIALS AND METHODS
2.1 Materials
For column experiments, silica sand with a low uniformity
coefficient of 1.25 was selected in order to simulate a sandy
aquifer having a relatively high hydraulic conductivity. Silica
sand of interest has 0.050 cm and 2.68 g/cm3 of physicalproperties such as the mean particle size and the particle
density, respectively. A ZVI powder in hand warmer and acommercial ZVI powder with a high reduction property wereused to compare the degree of nitrate reduction andretardation and were referred to as Material 1 and Material 2,
respectively.
2.2 Column experiments
In column experiments, ZVI powder of concern was mixed
with soil in order to examine the relation between the amount275

of ammonium ion was determined by the indophenol blue
method using a spectrophotometer (HACH, DR-4000).
To investigate the impact of water content on the nitrate
behavior, column experiments under unsaturated conditions
were carried out in a similar manner but a different pulse
volume of 20 cm3. As shown in Fig. 1(b) in unsaturated
experiments, instead of constant head reservoir, suction wasapplied at the bottom of the column to keep the moisture
content inside the column uniform while water was sprinkled
over the top of the medium until the outflow from the bottomof the column equaled the volume of the water input. Prior toan application of solute pulse, samples were adjusted tosteady state water flow condition.
2.3 Temporal moment analysis
Consider a sorbing and degradable chemical moving througha homogeneous medium at a steady, uniform, flow rate.
Assuming linear, equilibrium sorption and first-order kinetic
degradation for the solute, the governing transport equation
can be written as

Fig. 1 Schematic diagram of experimental column system

of ZVI powder and the degree of reduction of nitrate. Fromthe practical viewpoint, mixture ratios of ZVI powder were
set to 0%, 7.5%, 15% and 30% by weight. These soils withZVI powder were completely saturated before packing to
avoid the entrance of air and were filled into the column,
which has 30 cm in length and 5 cm in diameter, inincrements of 2.5 cm. Each layer was compacted to adjust thedry density prior to filling the next layer. The porosity can be
estimated in each experiment indirectly from themeasurements of the particle density and the dry soil bulk
density, resulting in approximately 0.41 of the porosity inflow field for both materials. In Fig. 1, the detailed column
design is illustrated.
After packing, as shown in Fig. 1(a), water was applied to thecolumn up to a specific level controlled by constant headreservoirs at the top and bottom of the saturated media, while
maintaining the saturated condition of porous media. Steadysaturated flow field was established in the column whenfluctuations in the observed drainage rate from the bottom
reservoir became negligible. A volume of 40 cm3 KNO3
solution under saturated condition was applied to the top of
the column to produce a pulse input with an initialconcentration of 1.5  10-4 g/cm3 of NO3-N. Prescribed solutesource condition was designed to clarify the transientconcentrations of NO3-N and to effectively provide theparameter identification relevant to solute transport using
temporal moment approaches, which will be stated in thefollowing subsection. Pore water samples at the end of thecolumn were taken at specific intervals using a fraction
collector. Nitrate ion in the pore water samples was analyzed
by capillary electrophoresis (G1600A, Agilent technologies,
USA). The basic anion buffer and a fused silica capillary with104 cm in length and 50 m internal diameter were obtained
from Agilent technologies. The temperature controlled
cartridge for fused silica was set at 15 °C. The concentration

c
 2c
c
R
  Lv p 2  v p
 c
x
t
x

(1)

where c is the concentration of solute, x is the coordinate, t is
the time, R is the retardation factor estimate, vp is the averagepore water velocity, L is the dispersivity, and is thedegradation rate [17].
Temporal moment analysis is convenient to use, as there is noneed to solve the transport model in real time and degrees of
spreading and asymmetry [18]. The n-th order temporal
moments are defined as

M n   t nc( x, t )dt0(2)

The nomalized temporal moments n at a location x, aredefined asn
M n 0 t c( x, t )dt
n 
 
M0
c( x, t )dt

0

(3)

Equations (2) and (3) are used to obtain experimental
temporal moments from concentration BTCs. Das andKluitenberg [17] derived theoretical first temporal momentusing Laplace formation of Eq. (1). By setting theexperimentally determined moments equal to the theoretical
moments, we can estimate the dispersivity and the retardation
factor [19].

276

L 

 p 2
2 12

(4)



 1 
ln

1  MRF 
1(7)

3 RESULTS AND DISCUSSION
3.1 Effect of ZVI powder on nitrate retardation
Representative BTCs under different mixture ratios of ZVI
powder as a function of pore volume for Materials 1 and 2 areshown in Fig. 2 and Fig. 3, respectively. Pore volume iscalculated as
Fig. 2 Representative BTCs for different mixture
ratios of ZVI powder in hand warmer under saturated
condition

Pore volume Fig. 3 Representative BTCs for different mixture
ratios of commercial ZVI powder under saturated
conditionR( 1  0.5t0 ) v 2p  4 Lv p

x(5)

where p is the distance between the source and theobservation location and t0 is the pulse duration. In all
experiments, t0 was set to 30 seconds.
The mass fraction of solute that is collected in the effluent
during column experiment is summed up to provide anestimate of the mass recovery fraction (MRF) [17].
 Q( L, t )c( L, t )dt
 Q(0, t )c(0, t )dt

MRF  0

(6)0where Q is the volumetric flow rates and L is the length of thesoil column. MRF allows to obtain the degradation rate [17].

qwt v p t

L
L

(8)

where qw is the water flux and  is the volumetric watercontent. When solutes travel without the adsorption onto thesoil particle, or the retardation, the pore volume becomes theunity and increases with the increase of the degree ofretardation.
As shown in Fig. 2 and Fig. 3, observed BTCs exhibit theshift of the peak concentration at the corresponding porevolume toward a larger pore volume with the increase of theamount of ZVI powder. A slight tailing of the BTCs can also
be seen in Fig. 2 and Fig. 3, except for the case of 0% of the
mixture ratio. These results imply that the influence of a
mixture of ZVI powder and soil, in a practical situation,
degradation of nitrate concentration is expected due to thereaction of nitrate with ZVI powder as a travel time of nitratefrom ground surface to groundwater becomes longer. Thus,
for the purpose of the retardation and the reduction of nitrate,
the use of ZVI powder might be effective in soils.
In order to quantify the degree of adsorption of nitrate onto
ZVI powder, the retardation factor is employed as a measureand is estimated based on Eq. (5). The relation between the
mixture ratio of ZVI powder and the retardation factor
estimates for Materials 1 and 2 are shown in Fig. 4 and Fig. 5,
respectively. The retardation factor estimates tend to increase
with increasing the percentage of ZVI powder and are within
the range of approximately 1.3 to 1.6 under saturated
condition. It is indicated that the opportunity of adsorption of
nitrate to ZVI powder increases as the amount of ZVI powder
becomes larger. On the other hand, under unsaturated
condition, the retardation factor estimates range fromapproximately 1.0 up to 1.2, indicating that the degree ofadsorption under unsaturated condition is lower than that
under saturated condition. This is because a less water content
and a higher pore velocity may lead to decrease the course oftransport under unsaturated condition in comparison withsolute pathways under saturated condition.

277

Fig. 4 Relation between the mixture rario of ZVI powder
in hand warmer and the retardation factor estimatesFig. 6 Relation between the mixture rario of ZVI powder
in hand warmer and the degradation rate

Fig. 5 Relation between the mixture rario of commercial
ZVI powder and the retardation factor estimatesFig. 7 Relation between the mixture rario of commercial
ZVI powder and the degradation rate

The values of retardation factor between two kinds of ZVI
powder, as seen in Fig. 4 and Fig. 5, vary with a similartendency. In addition, the retardation factor estimates for bothmaterials are almost the same values corresponding to thesame mixture ratio of ZVI powder. These results show that
there is no difference in the degree of adsorption between twomaterials.

seen in Fig. 4 and Fig. 5, retardation factor estimates are littlechange in the range of approximately 1.0 to 1.2 regardless ofthe mixture ratio of ZVI powder under unsaturated condition.
However, the degradation rates exhibit a little difference
between saturated and unsaturated conditions at the sameamount of ZVI powder as seen in Fig. 6 and Fig. 7. This may
attribute to the presence of air under unsaturated condition,enhancing a prompt oxidation of ZVI powder, since corrosion
process may induce the nitrate reduction [15].
As for the case of Material 1, the degradation rate estimates
are in the range of approximately 0.09 hr-1 to 0.15 hr-1 in Fig. 6,
when pH ranges from 7 to 9 in this study. The results are ingood agreement with a previous study. Alowitz and Scherer[9] reported the degradation rate of approximately 0.13 hr-1 at
pH 8. In the comparison with two kinds of ZVI powder,although the retardation factor estimates have similar values,
the degradation rates hinge on the material of interest. In thecase of Material 2 as shown in Fig. 7, the estimate ofdegradation rate at 7.5% of the mixture ratio is about 2 times
larger than that of Material 1 in Fig. 6. At 30% of the mixture
ratio, the degradation rate of Material 2 becomes about 4times larger than that of Material 1. These results show that

3.2 Effect of ZVI powder on nitrate reduction
To quantify the effect of ZVI powder on nitrate reduction
during a transport, degradation rate estimates are plotted as afunction of mixture ratio of ZVI powder for Materials 1 and 2
in Fig. 6 and Fig. 7, respectively. Regardless of water content,
the degradation rate increases as the amount of ZVI powder
increases for both materials, suggesting that the increase of
the amount of ZVI powder is expected to contribute nitrate
reduction. As stated in the preceding section, under saturated
condition, the opportunity of adsorption of nitrate to ZVI
powder increases with increasing the amount of ZVI powders
and may enhance the nitrate reduction in the duration
occurring adsorption phenomena. On the contrary to this, as278

Table 1 MRF by mixture ratio of ZVI powder
Nitrogenspecies

Condition ofwater content

NO3-N
NH4-N
NO3-N
NH4-N

Saturated
Saturated
Unsaturated
Unsaturated

0
1.000
0.000
1.000
0.000

Mixture ratio of ZVI powder (%)
Material 1
15
30
7.5
0.924
0.930
0.939
0.000
0.005
0.038
0.985
0.978
0.992
0.009
0.000
0.010

7.5
0.976
0.003
0.990
0.003

Material 2
15
0.853
0.044
0.951
0.048

30
0.795
0.090
0.893
0.109

commercial ZVI powder is more effective than ZVI powder
in hand warmer from the point of view of nitrate reduction.
3.3 Mass recovery fraction
In the process of nitrate reduction, the product of ammonium
and relevant mass recovery are of important. Table 1 lists the
MRFs for different mixture ratios of ZVI powder estimatedfrom Eq. (6) in order to compare the total amount of NO3-N
leaching from the bottom of the column with the total amountof NH4-N yielded by NO3-N reduction. The reaction of nitrate
reduction can be described by Eq. (9) [16].

NO 3-  10H   4Fe 0  NH 4  3H 2 O  4Fe 2 (9)As seen in Table 1, the MRF of NO3-N decreases with theincrease of the mixture ratio of ZVI powder, resulting in theincrease of the MRF of NH4-N, particularly in Material 2.
Interestingly, the loss of NO3-N does not necessarily
correspond to the production of NH4-N, especially in the
media mixed with Material 1. It appears that a part of nitrateor ammonium ion may strongly adsorb to ZVI powder andremain in the column, consequently leading to the unbalanceof the total MRFs of NO3-N and NH4-N.
Moreover, the MRF of NH4-N of Material 2 is about 10 times
larger than that of Material 1 at the same mixture ratio,
indicating the property of commercial ZVI powder in terms ofnitrate reduction to ammonium. On the contrary to this,nitrate is slightly converted to ammonium by the reaction
between nitrate and Material 1. Hence, single application ofMaterial 1 becomes less effective in the nitrate degradation.

Fig. 8 Relation between the mixture ratio of ZVI powder
in hand warmer and the dispersivity

3.4 Effect of ZVI powder on dispersivity

Fig. 9 Relation between the mixture ratio of commercial
ZVI powder and the dispersivity

The spread of the nitrate in soils is also important in a field
application of ZVI powder. The dispersivities of nitrate arealso estimated from Eq. (4), and are shown as a function of
mixture ratio of ZVI powder for Materials 1 and 2 in Fig. 8
and Fig. 9, respectively. As seen in Fig. 8 and Fig. 9, thelongitudinal dispersivities range from approximately 0.16 cm
to 0.70 cm and slightly increase with increasing the mixture
ratio of ZVI powder because of the influence of the mixture of
ZVI powder and soil. The longitudinal dispersivity at
maximum mixture ratio of ZVI powder is about 3 times larger
than the case without ZVI in porous media. Therefore, theeffect of the mixture of ZVI powders in field soils on nitratedispersion may be relatively lower.
4 CONCLUSIONSThis paper assessed the potential of zero-valent iron (ZVI)
powders as a nitrate reduction and retardation material inporous media. Under saturated and unsaturated flow
conditions, column experiments were conducted in silica sandwith a mixture of two kinds of ZVI powders. Experimental
breakthrough curves were analyzed using temporal moments279

not only to estimate the retardation factor and the degradation
rate but to quantify mass recovery fractions associated withnitrate and ammonium ions. The findings of this study are
drawn as follows:1)

The retardation factor estimates under saturated
condition were slightly larger than those underunsaturated condition regardless of the kinds of ZVI
powders.

2)

The increase of the amount of ZVI powder resulted in
the increase of the degradation rate regardless of thewater content. The degradation rates of commercial ZVI
powder were larger than those of ZVI powder in hand
warmer, indicating that commercial ZVI powder is more
effective in the reduction of nitrate to ammonium.

3)

According to the results of the mass recovery fraction, it
was confirmed that ammonium was produced by
reduction of nitrate in the presence of ZVI. The degreeof production of ammonium increased with increasing
the amount of ZVI powder, especially for commercial
ZVI powder.

4)

The longitudinal dispersivities slightly increased withincreasing the mixture ratio of ZVI powder.

[10] Maeda M, Ihara H and Ota T, “Deep-soil adsorption of nitrate in a
Japanese Andisol in response to different nitrogen sources,” Soil Sci.
Soc. Am. J., 72(3), 2008, pp. 702–710.
[11] Inoue K, Ihara I, Yoshino A and Tanaka T, “Assessment of the use ofhand warmer for nitrate retardation in porous media,” J. Water Environ.
Technol., 8(4), 2010, pp. 355–362.
[12] Jeen S-W, Gillham RW and Przepiora A, “Predictions of long-term
performance of granular iron permeable reactive barriers: field-scaleevaluation,” Environ. Sci. Technol., 23(6), 1989, pp. 630–636.
[13] Ruangchainikom C, Liao C-H, Anotai J and Lee M-T, “Characteristicsof nitrate reduction by zero-valent iron powder in the recirculated andCO2-bubbled system,” Water Res., 40, 2006, pp. 195–204.
[14] Murphy AT, “Chemical removal of nitrate from water,” Nature, 350,
1991, pp. 223–225.
[15] Cheng IF, Muftikian R, Fernando Q and Korte N, “Reduction of nitrate
to ammonia by zero-valent iron,” Chemosphere, 35(11), 1997, pp.
2689–2695.
[16] Westerhoff P and James J, “Nitrate removal in zero-valent iron packed
columns,” Water Res., 37, 2003, pp. 1818–1830.
[17] Das BS and Kluitenberg GJ, “Moment analysis to estimate degradation
rate constants from leaching experiments,” Soil Sci. Soc. Am. J., 60,
1996, pp. 1724–1731.
[18] Valocchi AJ, “Validity of the local equilibrium assumption formodeling sorbing solute transport through homogeneous soils,” Water
Resour. Res., 21(6), 1985, pp. 808–820.
[19] Pang L, Goltz M and Close M, “Application of the method of temporal
moments to interpret solute transport with sorption and degradation,” J.
Contam. Hydrol., 60(1-2), 2003, pp. 123–134.

5 ACKNOWLEDGMENT
The work reported here was supported by the Nippon Life
Insurance Foundation.
6 REFERENCES
[1]

[2]

[3]

[4]

[5]

[6]

[7]

[8]
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Sandor J, Kiss I, Farkas O and Ember I, “Association between gastriccancer mortality and nitrate content of drinking water: ecological study
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443–447.
Jansen DM, Buijze ST and Boogaard HL, “Ex-ante assessment of costs
for reducing nitrate leaching from agricultural-dominated regions,”
Environ. Monitor. Software, 14, 1999, pp. 549–565.
Schröder JJ, Aarts HFM, van Middelkoop JC, Schils RLM, Velthof GL,
Fraters B and Willems WJ, “Permissible manure and fertilizer use indairy farming systems on sandy soils in The Netherlands to comply
with the Nitrates Directive target,” Euro. J. Agronomy, 27, 2007, pp.
102–114.
Bennet P, He F, Zhao D, Aiken B and Feldman L, “In situ testing of
metallic iron nanoparticle mobility and reactivity in a shallow granularaquifer,” J. Contam. Hydrol. 116, 2010, pp. 35–46.
Mackay DM and Cherry JA, “Groundwater contamination:
pump-and-treat remediation,” Environ. Sci. Technol., 23(6), 1989, pp.
630–636.
Travis CC and Doty CB, “Can contaminated aquifers at Superfund sites
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280

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Using Spatial Moments in Conjunction with Image Processing to Estimate
Macrodispersion in Stratified Porous Formations
Suzuki K., Inoue K. and Tanaka T., Graduate School of Agricultural Science, Kobe University
Kobayashi A., Faculty of Environmental and Urban Engineering, Kansai University

ABSTRACT: Laboratory dye tracer experiments with a pulse source
were conducted under saturated unidirectional flow conditions in a
two-dimensional and vertically placed water tank. In tracer
experiments, homogeneous and stratified porous formations, whichwere comprised by a few combinations with three types of soil
particles, were of concern to examine the effect of variation of pore
structure on macrodispersion phenomena. A new methodology
using spatial moment analysis linked with image processing of a dye
tracer behavior was developed to estimate macrodispersivities bothin longitudinal and lateral directions. These results demonstrated
that the ratio of the longitudinal macrodispersivity to thelongitudinal microdispersivity ranged from about 2 to about 40.
Moreover, it is indicated that the layering of stratified porous
formations had an effect on the degree of longitudinal and transverse
macrodispersions.

Keywords: macrodispersion, stratified porous media, image
processing, spatial moment.
1. INTRODUCTION
The movement of groundwater in porous media is subject toconvection and dispersion, independently of any material
being transported [1]. Advection and forced-advection are
due to bulk movement of groundwater caused by regional
movement in the aquifer and by some man-made disturbancesuch as pumping wells, respectively, while natural convection
is relevant to differences in density. Dispersion results fromthe irregular movement of water in porous formations where
tortuosity of flow paths is induced. On a larger scale, theseirregularities are due to the presence of zones of different
hydraulic conductivities. It is progressively accepted that
dispersion of solutes by groundwater is governed bylarge-scale spatial heterogeneity of natural formations [2],[3].
The scale effect in solute transport through porous media isstrongly associated with the heterogeneity and has been the
topic of significant theoretical and numerical experiments. As
for layered media, Güven et al [4] conducted deterministic
analysis of dispersion in a perfectly stratified aquifer of finite
thickness and suggested that the time-dependent dispersivity
approached an asymptotic limit in time. Black and Freyberg[5] used Monte Carlo simulations to estimate the impact ofhydraulic conductivity uncertainty on concentration
uncertainty in a perfectly stratified aquifer. Fadili et al [6]employed a Lagrangian particle tracking to investigate thebehavior of concentration and statistical moments of thetransported tracer plume in multilayered media. The scaleeffect in solute transport in a stratified aquifer has also beenstudied under field conditions. Sudicky et al [7] observed thetransitional values of both longitudinal and transverse
dispersivities. Güven et al [8] demonstrated that the results ofa single-well tracer test could be interpreted without ascale-dependent dispersivity.
Although significant work has been performed through field
efforts, relatively few controlled laboratory efforts have beenreported. Recently, a few of the dye tracer experiments
combined with an image processing technique have beenstudied to quantify the behavior of dye tracers [9],[10], whichhave been used for a long time to trace water flow and solute
movement in aquifers. McNeil et al [11] performedintermediate-scale laboratory experiments in heterogeneous
porous media and characterized the spatial distribution ofsolute concentrations using image analysis methods,
providing concentration distributions and longitudinal
dispersivity in a corresponding porous formation. However,thevariationoflongitudinalandtransverse
macrodispersivities corresponding to the heterogeneity was
not described.
The objectives of this study are to investigate the transport
behavior of dye tracer in homogeneous and stratified porous
formations under saturated conditions, to estimate transportparameters including the longitudinal and transverse
dispersivities and dispersion coefficients and to assess
macrodispersion phenomena. An image processing technique
is used in conjunction with spatial moments for the reliable
parameter estimation and is applied to a flow field with alayered formation.
2. TRACER EXPERIMENTS
2.1 Dye tracers and experimental apparatus
Dye tracing is widely used to characterize water flow andsolute transport behavior in porous media. Previous work
demonstrated that colored dye tracers could be successfully
used to monitor solute transport in a porous medium confinedin a transparent container [12]. In this study, a dye tracer,
Brilliant Blue FCF, with the initial concentration of
0.4 mg/cm3 is employed. The specific gravity of dye tracer is
1.0001 measured using the specific gravity meter and can be
therefore regarded as almost the same as that of water.
Although the initial concentration of dye tracer is determined
to be low enough to avoid density-induced flow effects, there
is no denying that the effect of gravity on solute transport
during the course of movement.
Tracer experiments are carried out in a two-dimensional andvertically placed water tank with the dimensions of 62 cm

281

Table 1 Properties of soil materials.

Mean particle size (cm)Uniformity coefficient (-)
Hydraulic conductivity (cm/s)
Dry density (g/cm3)

K4
0.085
1.80
0.751
2.68

Material
K5
0.050
1.25
0.268
2.68

K6
0.030
1.31
0.0571
2.68

Table 2 Experimental cases.
Case

Layer number
2
3
4
K4
K5
K6
K4
K5
K4
K6
K4
K5
K6
K4
K5
K4
K4
K6
K4
K5
K4
K5
K6
K4
K6
1

K4
K5
K6
K4-5
K4-6
K4-5-6
K4-5-4K4-6-4K5-4-5K6-4-6

Fig.1 Schematic diagram of experimental apparatus.
width, 50 cm height and 1 cm thickness. The water flow tankcontains silica sand to form transparent quasi
two-dimensional solute transport phenomena and consists of
two acrylic plates with 2 cm thickness. Schematic diagramof experimental apparatus is shown in Fig.1. Constant headwater reservoirs connected to the upstream and downstream
ends of the water tank are used to control the hydraulicgradient. One transparent acrylic plate allows for theobservation of migrating dye tracer and for the measurementof the piezometric head at 20 observation points using
manometers.
In tracer experiments, in order to elucidate the effect of thelayering on the degree of macrodispersion, ten stratified
porous formations are of interest and are comprised using
three types of soil materials. Physical properties of soilmaterials, referred to as K4, K5 and K6, are shown in Table 1.
For the base case, each soil material is filled in all Layers 1
through 4 shown in Fig.1 and comprises a homogeneous
porous formation, providing “Single type formation” as
Cases K4, K5 and K6. In two-layered porous formations,
Layers 1 and 2 are filled with soil material K4 as the upper
layer, while soil material K5 or K6 comprise the lower layer
including Layers 3 and 4. As Cases K4-5 and K4-6, these two
types of porous formations are referred to as “Step typeformation”. As one possible Step type formation, all threematerials comprise flow fields where soil materials K4, K5
and K6 are filled with Layer 1, Layer 2 and 3, and Layer 4,
respectively. In the same manner except for the use of two
types of soil materials, soil material K4 comprises the top andbottom layers and the middle layer consists of material K5 or
K6, referred to as “Convex type formation” and as Cases
K4-5-4 and K4-6-4. In Convex type formation, soil materialis switched to another material in order to create two different
flow fields as “Concave type formation”. In Table 2,experimental cases and corresponding layering formations

Type of stratified
porous formation
SingleSingleSingle
Step
Step
Step
ConvexConvexConcave
Concaveusing three soil materials are summarized as well as theassigned name of stratified porous formations of concern.
2.2 Experimental procedure
Materials are completely washed and saturated before
packing to remove organic chemicals attached to the particle
surface, to avoid entering the air and to conduct experiments
under the saturated condition. In the process of creation ofhomogeneous and heterogeneous flow field formations, waterflow tank is filled with water and material of interest frombottom to top in 5 cm layers to achieve uniform packing. In
this process, saturated material is funneled using an extended
funnel. Each layer is compacted prior to filling the next layer,
resulting in 0.41 ± 0.02 of the porosity for all materials. The
porosity of each material can be estimated indirectly frommeasurements of the particle density and the dry soil bulk
density.
After packing, water is applied to the flow tank under aspecific hydraulic gradient controlled by constant head water
reservoirs at the upstream and downstream sides, while
maintaing saturated condition of porous media. A steadysaturated flow field is established in the flow tank whenfluctuations in the observed drainage rate, which is effluent
from the constant head water reservoir, and piezometerreadings can become negligible.
After reaching steady state flow conditions, in order tosimulate a pulse type source, dye tracer with the volume of
5 cm3, which makes flow paths visible, is uniformly injectedalong the whole thickness of the flow tank. Tracer injection
requires approximately 20 seconds so as not to possibly
induce complications in the flow field, resulting in cylindricalsource having a height of 1 cm and a radius of approximately
1.2 cm. In an injection system, a needle is inserted through theinjection port with 0.05 cm of the radius on the face of acrylic282

Fig.3 Relation between the image brightness and the dye
tracer concentration.

Fig.2 Image examples of dye tracer transport.
plate in order to create a two-dimensional transport state.
During the experiment, the profiles of tracer migration are
periodically recorded using a digital camera. The distance
between the digital camera and the experimental flow tank is
set at approximately 40 cm. A time series of images can thenbe processed and analyzed through image analysis forparameter estimation. Several experimental cases arerepeatedly conducted under various hydraulic gradientconditions within approximately one order range of Reynolds
number.
2.3 Transport process and image calibration
Fig.2 exhibits some images associated with dye tracer
distributions in porous media. Data recorded by the digital
camera successfully indicate different intensities in dye tracer
distributions, suggesting different concentrations of the dye
tracer. Moreover, the tracer experiments may be regarded as
effectively two-dimensional since dye tracer is injectedacross the full 1 cm thickness of the flow tank.
In order to establish the relationship between the image
intensity of a pixel and dye tracer concentration, a calibrationis conducted. Under identical experimental conditions, aknown concentration of dye tracer is injected into a
corresponding porous formation without a hydraulic gradient.
The spread of dye is captured by the digital camera. The sameprocedure is repeated using different concentrations of dye
tracer, providing the relation between the image intensity anddye tracer concentration as shown in Fig.3 where calibration
formulas are also shown and are specific to the experimental
configuration. The concentration of the dye is proportional tothe pixel brightness over the range of 0 mg/cm3 to 0.4 mg/cm3
as seen in Fig.3.

2.4 Spatial moment approach
A commonly used measure of dilution is the spatial momentsof the concentration spatial moments of aqueousconcentrations distributed in space in porous media and are
calculated from snapshots of tracer plume at given times asfollows [13].

M ij (t )   C ( x, z, t ) x z dxdz
i

j

(1)

 

where x and z are the Cartesian coordinates, C is the soluteconcentration, t is the time, Mij is the spatial moments
associated with the distribution of tracer plume at a certain
time, and i and j are the spatial order in the x and z coordinates,
respectively.
The tracer brightness distribution can be converted to aconcentration distribution by the calibration, providing ananalogy between Eq.(1) and Eq.(2).

M ij (t )   H ( x, z) B( x, z, t ) x z dxdz

 

i

j

(2)

where H(x,z) is the area per unit pixel and B(x,z,t) is thebrightness at a corresponding pixel.
The centroid of plume concentration distribution is calculated
as the normalized first order spatial moment by the following
equation.

xc 

M 10
M
, zc  01
M 00
M 00

(3)

where xc and zc are the cetroid locations of plumeconcentration distribution in the x and z coordinates,
respectively. The liquid phase second order spatial moments
are also computed as follows.

283 M 20
 xc2

 σ xx  xz   M 00
 
σ ij    zx  zz   M 11  zc xc
M
 00

M 11

 xc zc 
M 00

M 02

 zc2 
M 00


(4)

where ij is the second order spatial moments.
Longitudinal and transverse macrodispersivities from spatial
moments of the distributed tracer plume are calculated as [10]AL 

1  xx
,
2 c

AT 

1  zz2 c(5)where AL is the longitudinal macrodispersivity, AT is thetransverse macrodipersivitity and c is the travel distance ofthe center of tracer plume in the mean flow direction at time t .
Furthermore, the longitudinal and transverse dispersion
coefficients are estimated based on the relation between theflow velocity of dye tracer and the dispersivity as thefollowing equations.

DL  v AL , DT  v AT

(6)

where DL is the longitudinal dispersion coefficient and DT is
the transverse dispersion coefficient.
An advantage of spatial moment analysis in conjunction with
an image technique is that the underlying physical model is
not needed unlike other techniques such as fitting advection
and dispersion equation. This point may lead to the cost
reduction related to labor sampling of other conservative
chemicals such as NaCl and KBr.
3 RESULTS AND DISCUSSION
3.1 Microdispersion coefficient estimates
Estimates of longitudinal and transverse microdispersion
coefficients in single layer formations of Cases K4, K5 andK6 are shown in Fig.4 as a function of Reynolds number,which is expressed as follows [1]

Re 

vd50



(7)

where Re is the Reynolds number, v is the seepage velocity,
d50 is the mean particle size and is the kinematic viscosity.
Harleman and Rumer suggested that a nonlinear relation
between dispersion coefficients and Reynolds number asfollows [14] .

DL /  bL Re f L , DT /  bT Re fT(8)

where bL, bT, fL and fT are constants, which depend on particle
shape. The best fittings based on Eq.(8) with the empiricalFig.4 Variation of normalized longitudinal and transverse
dispersion coefficients as a function of Reynolds number.
constants of 2.48 of bL, 0.66 of fL, 0.257 of bT and 0.53 of fT. In
this figure, empirical relations suggested by Harleman andRumer [14] are also depicted for the normalized longitudinal
dispersions with the constants 0.66 of bL and 1.20 of fL and forthe transverse dispersions with the constants 0.036 of bT and
0.70 of fT. Both longitudinal and transverse dispersion
coefficients increase not only with the Reynolds number inflow fields but with the mean particle size. This tendency isquite natural in solute transport phenomena in homogeneous
porous media [1], indicating the validity of the methodology
applied herein. On the other hand, the fitting curves obtainedin this study provide larger values compared to the empirical
relations because the difference between physical properties
including the uniformity coefficient and the mean grain sizeof materials in tracer experiments employed in this study andother literature may affect a porous formation, and result in awide variety of tortuous pathways during the course of tracer
migration.
3.2 Macrodispersivity estimates
As stratified porous formations of Step, Convex and Concave
type formations, as well as Single type formation,
longitudinal and transverse macrodipersivites in eachformation are plotted in Fig.5 as a function of Reynolds
number. Longitudinal macrodispersivites in multilayered
porous media are larger than those in single layers of Cases
K4, K5 and K6. This is attributed to the distribution of
hydraulic conductivity comprising flow fields and is one
evidence occurring macrodispersion phenomena.
As for Step type formations, longitudinal macrodispersivities
in Case K4-6 are larger than those in Case K4-5. Moreover,longitudinal macrodispersivites in Case K4-5-6 are lower
values than those in Case K4-6. It is inferred that larger
difference of hydraulic conductivity between neighboringlayers results in larger values of longitudinalmacrodispersivity. On the other hand, the difference oftransverse macrodispersivities is relatively small despite of
the stratified porous formation. This is because the flow
direction is parallel to the layers and the vertical component284

middle layer in three layers present a unique tendency as
shown in Fig.7. Whereas a little difference of the longitudinalmacrodispersivity estimates between Cases K5-4-5 andK6-4-6 appears, transverse macrodispersivity estimates
depend on the soil materials comprising the top and bottom
layers. A few studies [15], [16] mentioned a solute transfer
across a discontinuity such as a layered formation andsuggested a probability expression as follows

P1 
Fig.5 Variation of longitudinal and transverse
macrodispersivities as a function of Reynolds number in
Step type formation.

D1
D1  D2, P2  1  P1 

D2
D1  D2(9)

where D1 is the transverse dispersion coefficient in a layer inwhich tracer exists, D2 is the transverse dispersion coefficientin the neighbor layer, P1 is the transfer probability that a
tracer remains in the same layer in which tracer exists and P2
is the transfer probability that a tracer goes into theneighboring layer.
As seen in Fig.4, Case K4 has the largest value of transverse
macrodispersion coefficient among flow fields of Single typeformation. Therefore, according to the expression of Eq.(9),tracers migrating in the K5 or K6 layer may tend to travel
toward the K4 layer. Hence, as shown in Fig.7, in Concave
type porous formations employed herein, it appears that thedifference of hydraulic conductivity between layers does notaffect on the variation of longitudinal macrodispersivity
estimates but on transverse macrodispersivity estimates.

Fig.6 Variation of longitudinal and transverse
macrodispersivities as a function of Reynolds number inConvex type formation.

Fig.7 Variation of longitudinal and transverse
macrodispersivities as a function of Reynolds number in
Concave type formation.
of flow velocity is almost zero in this experimental field. As
for Convex type formations, the same tendency as the resultsof Step type formation appears for longitudinal and transverse
macrodispersivity estimates as seen in Fig.6.
On the other hand, unlike the results of Step and Convex typeformations, Concave type formations where the soil material
having the largest hydraulic conductivity comprises the

3.3 Macrodispersivity vs. microdispersivity
As for results in stratified porous formations, Fig.8 exhibits
the variation of the transverse macrodispersion coefficient as
a function of Reynolds number. Additionally, for the purposeof comparison between macrodispersion and microdispersion
coefficient estimates, transverse dispersion coefficients in
Single type formation in Cases K4, K5 and K6 are alsoplotted. Transverse macrodispersion coefficient estimates in
Step and Convex type formations are larger than thoseestimated in single layers K4, K5 and K6. This is because
tracers migrating in K5 or K6 layer may tend to flow into theK4 layer as aforementioned above. Opposite to this result,transverse macrodispersion coefficient estimates in the
Concave type formation of Case K6-4-6 are smaller thanthose estimates in single layers K4, K5 and K6. This also
attributes to the difference of hydraulic conductivity between
two layers. In Concave type formations, tracers located in the
top and bottom layers have a certain possibility to flow into
the middle layer, which has the largest hydraulic conductivity.
Therefore, the degree of transverse dispersion may tend to
decrease as tracers move. However, as shown in Fig.8, one of
the Concave type formations, Case K5-4-5, leads to almost
the same values of transverse macrodispersion coefficients as
those in Single type formations. According to Eq.(9), thetransfer possibility of tracers depends on the transverse
dispersivity. There is a small difference between not only285

estimated based on spatial moments in conjunction withimage processing. The following findings have been clarified.
1. Image processing is a non-intrusive approach and candirectly characterize the solute movement in porous
media.
2. Transverse macrodispersion coefficient estimates in Step
and Convex type formations are larger than those inSingle type formations. Contrary tendency is seen in
Concave type formation due to the effect of the layering
comprising stratified formations.
3. Regardless of stratified porous formations, the ratio of thelongitudinal macrodispersivity to the longitudinalmicrodispersivity exceeds the unity, and ranges fromabout 2 to about 40.
4. The layering of stratified porous formations has an impact
on longitudinal and transverse macrodispersion.

Fig.8 Variation of normalized transverse dispersion
coefficients as a function of Reynolds number.

5 REFERENCES[1][2]

Fig.9 Variation of normalized longitudinalmacrodispersivities as a function of Reynolds number.
transverse dispersivities but the values of hydraulicconductivity in soil materials K4 and K5, resulting in similarvalues of Case K5-4-5 to the results of Single type formation.
Fig.9 demonstrates that the variation of the longitudinalmacrodispersivities divided by the microdispersivity(L=0.135cm), which is an estimate in Single type formationof Case K4, as a function of Reynolds number. Regardless ofstratified porous formations, the ratio of the longitudinalmacrodispersivity to the longitudinal microdispersivity
exceeds 1, and ranges from about 2 to about 40. Moreover,there is a marked difference between estimated values of
AL/L in stratified porous formations. In Step type formation
the values of AL/L are larger than those in Convex andConcave type formations due to the difference of the degreeof transverse dispersion induced from the layering
comprising stratified formations.
4 CONCLUSION

Bear J, Dynamics of fluids in porous media. Dover Publications.
Dagan G, “Time-dependent macrodispersion for solute transport inanisotropic heterogeneous aquifers,” Water Resour. Res., 24(9), 1988,
pp. 1491–1500.
[3] Gelhar LW, Welty C and Rehfeldt KR, “A critical review of data on
field-scale dispersion in aquifers,” Water Resour. Res., 28(7), 1991, pp.
1955–1974.
[4] Güven O, Moltz, JM and Melville JG, “An analysis of dispersion in a
stratified aquifer,” Water Resour. Res., 20(10), 1984, pp. 1337–1354.
[5] Black TC and Freyberg DL, “Stochastic modeling of verticallyaveraged concentration uncertainty in a perfectly stratified aquifer,”
Water Resour. Res., 23(6), 1987, pp. 997–1004.
[6] Fadili A, Abaou R and Lenormand R, “Dispersive particle transport:
identification of macroscale behavior in heterogeneous stratified
subsurface flows,” Math. Geol., 31(7), 1999, pp. 793–840.
[7] Sudicky EA, Cherry JA and Frind EO, “Migration of contaminants ingroundwater at a landfill: a case study 4. A natural-gradient dispersion
test,” J. Hydrol., 63, 1983, pp. 81–108.
[8] Güven O, Falta RW, Moltz, JM and Melville JG, “Analysis andinterpretation of single-well tracer test in stratified aquifers,” Water
Resour. Res., 21(5), 1985, pp. 676–684.
[9] Schincariol R and Schwartz FW, “An experimental investigation ofvariable density flow and mixing in homogeneous and heterogeneous
media,” Water Resour. Res., 26(10), 1990, pp. 2317–2329.
[10] Inoue K, Takenouti R, Kobayashi A, Suzuki K and Tanaka T,“Assessment of a UV excited fluorescent dye technique for estimating
solute dispersion in porous media,” J.
Rainwater
Catchment
Systems, 17(1), 2011, pp. 1–9.
[11] McNeil JD, Oldenborger GA and Schincariol RA, “Quantifying
imaging of contaminant distributions in heterogeneous porous media
laboratory experiments,” J. Contam. Hydrol., 84, 2006, pp. 36–54.
[12] Flury M and Flühler H, “Tracer Characteristics of Brilliant Blue FCF,”
Soil Sci. Soc. Am. J., 59(1), 1995, pp. 22–27.
[13] Tompson AFB and Gelher LW, “Numerical simulation of solutetransport in three-dimensional, randomly heterogeneous porous
media,” Water Resour. Res., 26(10), 1990, pp. 2541–2562.
[14] Harleman DRF and Rumer RR, “Longitudinal and lateral dispersion in
an isotropic porous media,” J. Fluid Mech., 16, 1963, pp. 385–394.
[15] Hoteit H, Mose R, Younes A, Lehmann F and Ackerer P,“Three-dimentional modeling of mass transfer in porous media using
the mixed hybrid finite elements and the random-walk methods,” Math.
Geol., 34(4), 2002, pp. 435–456.
[16] Uffink GJM, “A random-walk method for the simulation ofmacrodispersion in a stratified aquifer,” Proc. IAHS symposia, 65,
1985, pp. 26–34.

In the present study, the behavior of macrodispersion in
stratified porous formations has been assessed.
Macrodispersion coefficient and microdispersivity have been286

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Novel Technology of Desertification Control and Ecological Restoration
with an Hydrolysis Polyurethane(W-OH)
Weimin Gao1, Zhiren Wu2, Zhishen Wu3, Kentaro Iwashita4, Hirondo Inagaki2, Caiqian Yang5
1. Graduate School of Engineering, Ibaraki University, Hitachi, 3168511, Japan
2. JCK Co. Ltd., Anjo, Aichi, 446-0071 Japan
3. Department of Urban and Civil Engineering, Ibaraki University, Hitachi, 3168511 , Japan
4. Department of Urban and Civil Engineering, Meijo University, Nagoya, 468-8502, Japan
5. International Institute for Urban Systems Engineering, Southeast University, Nanjing, 211189, China


ABSTRACT: A novel hydrolysis polyurethane(W-OH) is an
eco-type solidification material that is used in desertification control.
This paper presents a new method for the ecological treatment ofdesertification using W-OH and discusses the large-scale production
technology of W-OH. Solidification of W-OH is by a cross-linking
reaction between W-OH and water in a broad proportion. The
reaction is rapid (2 to 20 min). The resulting material is elastic withlarge water holding properties. A sand-fixing layer can be formed byspraying an aqueous solution of W-OH on a sand surface. We findthat the resulting sand-fixing layer has good elasticity, suitable
porosity, excellent capability of holding water, heat and fertilizer,
and it can supply a stable and suitable environment for vegetation.
Moreover, the degradation period of W-OH can be controlled byadding different concentrations of a hindered amine light stabilizer(HALS or the W-US as product name). The degradation of thesand-fixing layer can be modified according to the local
environmental conditions. The degradation products are non-toxic
and do not present an environmental hazard. Construction of W-OH
desertification control is simple: it does not require large mechanicalequipment and is suitable for large-scale construction.

Key words: W-OH organic composite material; W-US;
Chemical-biological sand fixation; Desertification control;
Ecological restoration
1. INTRODUCTION
Currently, there are three major methods to combat
desertification [1]: the physical and mechanical method, thechemical method and the biological method. The physical andmechanical method consists of placing a sand trap, such as a
grass grid, on the sand surface. While the sand trap isbeneficial in the prevention of sand-shifting, it is also plaguedby many problems, such as the labor associated withinstallation, low efficiency in the prevention of desertification,and poor biological-environmental compatibility, among
others [2]–[4]. As the sand trap ages, the material degrades
and gradually rots. The trap loses its function when sandparticles begin to shift with the wind. The physical andmechanical method is thus not suitable for large-scale use andis not a viable solution for desertification [5], [6].
Chemical sand fixation is essentially a specializedmechanical sand-fixing method. Critical development ofchemical sand fixation methods has taken place over the past
80 years. With improvements in the chemical industry, many
new organic and inorganic sand-fixing materials have beendeveloped and applied in desertification control [7], [8].
However, this method has the same inherent problems asmechanical sand fixation: the sandy soil is only fixed for ashort term and thus does not provide a reliable solution todesertification. On exposure to UV radiation and wind/sanderosion, the materials gradually degrade and eventuallydisappear. Additional limitations include high costs andcomplex construction. Moreover, the fixed surface is
unsuitable for plants growth and typically harmful to theenvironment [9].
Biological sand fixation not only controls desertification but
also improves microclimate conditions, restores theenvironment and provides additional benefits in areas withrelatively high rainfall. Biological sand fixation is afundamental solution for the problem of desertification.
However, because of the poor climate and hydrologicalconditions often found in areas at risk for desertification, it is
impossible for plants to survive and grow. Therefore, it isnecessary to study a novel control technology for improving
the growth environment and survival rate of plants under
extreme desertification conditions [10], [11].
The development of chemical-biological integrated sandfixation technology is a current trend in the field ofdesertification control. However, it is difficult to combine
existing chemical sand-fixing materials with a biologicalsolution due to their poor durability and plant affinity.
Therefore, such methods have not been applied forlarge-scale desertification prevention and control. In thispaper we introduce a novel hydrophilic polyurethane(W-OH)
sand stabilizer. The material is prepared with ethylene oxide(EO) and propylene oxide (PO) random copolyethers, excess
modified 4, 4’-diphenylmethane diisocyanate and otherfunctional materials. Solidification is obtained by across-linking reaction between W-OH and water in a broad
proportion. The reaction is quick and is completed within therange of 2 to 20 min. The resulting solid is elastic, with greatwater-holding capability. A sand-fixing layer is formed byspraying W-OH aqueous solution on the sand surface. The
thickness of the layer is between 1and 5 cm, with a porosity
between 8 and 25%(by volume percent). The layer is robust
enough to prevent erosion caused by sand and wind. The
sand-fixing layer is characterized by good elasticity, suitable
porosity and excellent capability of preserving water, heatand fertilizer. Therefore, it provides a stable and goodenvironment for the growth of vegetation. The degradation
period can be controlled from 1 to 20 years by using different
concentrations of a hindered amine light stabilizer(HALS or
W-US as the product name). Therefore, the method of287

combating desertification with W-OH proves to be asustainable technique [12]–[15].
2. The characterization of W-OH
2.1 Physical and chemical characteristics of W-OH
W-OH is a light yellow, oily liquid. Water is used as asolidifying agent. W-OH is soluble in water at a broad rangeof concentrations. Solidification is achieved using across-linking reaction between W-OH and water. The
reaction is quick, ranging from 2 to 20 min. W-OH has manypotential applications, including use in soil stabilization andsand fixation, and as a waterproofing agent in irrigation
ditches or a dust proof layer. Recent research on soil
stabilization and sand fixation has resulted in manytechnological breakthroughs. In the work presented here, we
stress that W-OH has many excellent properties, such as ①
Excellent durability: the resulting solid is resistant to UV
irradiation for up to 20 years, and this property can becontrolled by adjusting the components of W-OH; ② Goodmechanical properties and high adhesive forces with othermaterials such as sand and soil; ③ Fast solidification with alarge range of water concentrations, and the resulting solid isvery elastic with great water holding capacity; ④ Good
resistance to acid and alkaline erosion, hydrolysis and salt
corrosion; ⑤ Is environmentally friendly.
2.2 Solidifying characteristics of W-OH
W-OH can be dissolved in water, emulsified and dispersed
homogeneously. The solidification proceeds over several
minutes following the introduction of water. As shown in Fig.
1, when the concentration of W-OH is less than 3%(by weightpercent), the cross-linking reaction forms a very weak solid
due to the low cross-linking density and low molecularweight of the solidified W-OH. Therefore, a concentration ofat least 3% of the sand-fixing material in water is necessary
for sand fixation. With an increase in temperature from 5 to
30℃, molecular motion is accelerated and the viscosity of thecross-linking reaction increases. However, this isaccompanied by a decrease in the cross-linking density and acorresponding decrease in the viscosity and the strength of thesolidified W-OH. By comparing the solidification conditions

16

pH7
3%
4%
5%

Solitification Time(min)

14
12
10
8
6
4
2
0
275

280

285

290

295

300

Temperature(K)Fig.1 Relationship between the solidifying time of W-OH andtemperature

at 5℃ and 30℃, we see that the time for solidification
decreases by more than 50%, from 19 min to 7 min at aconcentration of 3% sand-fixing agent, from 12 min to 2.2
min at a concentration of 4% sand-fixing agent, and from 8
min to 2 min for a concentration of 5% sand-fixing agent.
Therefore, the solidification time and final strength of thematerial is a function of reaction temperature and W-OH
concentration.
2.3 Safety of W-OH
W-OH uses water as solidifying agent. On completion of the
W-OH and water reaction, all of the W-OH present isconsumed. W-OH contains no heavy metal ions. The safety of
W-OH was measured and certified by public authorization,with respect to both plant and animal toxicity. The graph inFig. 2 shows the results of the W-OH fish toxicity test.
Results from both fish and mouse toxicity tests demonstratethat W-OH is an environmentally friendly material and does
not harm living organisms. This is a prerequisite for any
material used in desertification control.

Fig. 2 Security test
3. The mechanism and construction technology ofsustainable decertification control with W-OH
3.1 The mechanism of sustainable decertification control
with W-OH
The work presented here is a novel conception of sustainable
desertification control with W-OH and we aim to further
develop and study this technology. Sustainable desertification
control relies on a combination of chemical sand fixation andorganic biological sand fixation. Desertification control andecological restoration can be divided into two stages: first,
chemical sand fixation with W-OH, followed by biologicalsand fixation. The schematic diagram illustrating themechanism of sustainable sand fixation and vegetation using
W-OH is shown Fig. 3. W-OH is sprayed on a sand surface
that has been treated with seed and fertilizers. Several
minutes later, a sand-fixing layer is formed on the sandsurface and the objective of chemical sand fixation isachieved. However, chemical sand fixation only addresses
the stability of the soil and sand dune, and with chemical
fixation alone, it is impossible to achieve sustainablebiological sand fixation and complete ecological restoration.

288

The sand-fixing layer with W-OH has a thickness of morethan 1 cm and porosity of approximately 20%. This layer
presents a suitable environment for plant germination andgrowth. Thus, the second stage of sustainable biological sandfixation can be achieved in the presence of W-OH. The
degradation period of the W-OH sand-fixing layer can becontrolled according to the local environmental conditions. It
has been found that plants can germinate one week after
spraying W-OH aqueous solution, and the sand-fixing layer
degrades completely after 2 to 3 plant life cycles. In this way,
the natural transition between chemical sand fixation andbiological sand fixation is achieved.

Seed

Sand

Compostingmaterial

Fertilizer

remote desert. The construction technology with W-OH ismainly clarified into two stages. The first stage is landconsolidation which includes mixture of seed, compostingmaterial and fertilizers with sand, spraying the mixture on thesand surface, land grading and filling, followed by spraying
water. The second stage is spraying W-OH on the sandsurface with a special equipment of Y type. It contains two
pumps and tunnels for transporting W-OH and water
respectively, and W-OH and water mixes at nozzle and thenthe solution is sprayed on the sand surface. The concentrationis controlled by adjusting water pump.
The technology needed for desertification control with W-OH
consists solely of spraying equipment, which is suitable forlarge-scale application. The concentration of W-OH can becontrolled form 2% to 7%, according to the needs of a
specific application. This technology benefits from simple
equipment and construction processes.
4 Properties of the W-OH sand-fixing layer
4.1 Appearance of the sand-fixing layer
As a result of low viscosity prior to solidification, W-OH caneasily penetrate into the sand on spraying, and a sand-fixing
layer forms after few minutes. Fig. 4 shows the appearance ofthe sand-fixing layer. The sand-fixing layer is characterized
by good elasticity and sufficient porosity, which is necessary
for plant growth. Fig. 5 shows the layer thickness and initialporosity of the sand-fixing layer as a function of the amountof W-OH that is sprayed. If the amount of W-OH is too low,
the thickness of the sand-fixing layer will be too thin toachieve proper sand fixation. Therefore, the spraying quantity
is set to more than 3 L/m2. When the amount of W-OH
sprayed on the sand is 3 L/m2, the thickness of the sand-fixing
layer is approximately 1.4 cm with an initial porosity of 20%.
The sand-fixing layer not only fixes sand but is also suitablefor plant growth. The thickness of the sand-fixing layer
increases linearly with the quantity of W-OH that is sprayed.
However, with an increased quantity, the initial porosity
decreases due to the higher concentration of the sand-fixing
material in the sand, and the smaller pore size. When thequantity of W-OH sprayed on the sand is 10 L/m2, thethickness of the sand-fixing layer is approximately 3 cm, andthe porosity decreases to 8%. These results demonstrate thatthe thickness and porosity of the W-OH sand-fixing layer cana. Distribution of seed and fertilizers
Interface

Homogeneous
distribution of pores

Chemical sand fixationb. W-OH spraying-solidifying
Sand-fixing layer
Plant

Biological sand fixationc. Sustainable biological sand fixation and ecologicalrestoration
Fig. 3 The mechanism of sustainable sand fixation andvegetation with W-OH
3.2 The construction technology
The construction technology of sustainable decertification
control with W-OH is simple and convenient without any
large mechanical equipment. It is very important for
application in desertification control due to dismal and

289

Fig. 4 Appearance of the sand-fixing layer with W-OH

20
18
16
14

2.0

12
1.5

10

Initial Porosity%

2.5

22

8

1.0
2

3

4

5

6

7

8

9

10 11

Spraying Amount/(L/m2)

Fig. 5 Relationship between thickness and initial porosity ofthe sand-fixing layer and the spraying amount of W-OH
be controlled according to the needs of the local condition byadjusting the quantity of the W-OH solution sprayed.
4.2 Performance of compressive stress and wind erosion
resistance
The sand-fixing layer is an elastic material. When theconcentration of W-OH is increased, the molecular weight ofthe solidified W-OH network increases, the cross-linking
density increases, the intensity solidified W-OH increases,
and the adhesion between the solidified network and sandparticles is enhanced. Therefore, the compressive stress of thesand-fixing specimens increases linearly with increasing
concentration of W-OH. As the concentration increases from
3% to 7%, the compressive stress enhances from 0.37 MPa to
1.27 MPa (Fig. 6). The linear formula is shown in Eq. 1, with
R2 equal to 0.97. Therefore, the sand treated with W-OH has
adequate mechanical properties to withstand strong wind/sand erosion and an adverse desert environment.

4.3 Performance of durability and water retention
It is well known that a desert environment is characterized bystrong sunshine, UV radiation, low annual precipitation andhigh water evaporation. As a result, it is prerequisite for
W-OH sand-fixing material to have good durability and water
retention, especially under UV radiation. The durability of
W-OH was studied under accelerated aging test conditions
with UV radiation. Degradation was inhibited by the
inclusion of W-US. Fig. 7 shows the results of W-OH
durability tests using different concentrations of W-US. Tests
demonstrate that W-OH durability is significantly improvedby the inclusion of W-US. The decomposition rate of thesand-fixing layer is decreased by 80% when a W-US

1.4

1.0
0.8

OH3%
W-OH3%
OH3%+JF95:0.015%
W-OH3%+W-US0.015%
W-OH3%+W-US0.05%
OH3%+JF95:0.05%
OH3%+JF95:0.07%
W-OH3%+W-US0.07%
OH3%+JF95:0.1%
W-OH3%+W-US0.1%
OH3%+JF95:0.15%
W-OH3%+W-US0.15%
W-OH3%+W-US0.2%
OH3%+JF95:0.2%
OH3%+JF95:0.25%
W-OH3%+W-US0.25%
W-OH3%+W-US0.3%
OH3%+JF95:0.3%
W-OH3%+W-US0.35%
OH3%+JF95:0.35%

40
35
30
25
20

W-US: 0%15
10W-US: 0.015%

5

W-US: 0.05%

0
0

50

100

150

200
250
照射时间（hour）

300

350

400

450

Radiation time (hour)

0.6
0.4
0.2
0.0
2

3

4

5

6

7

8

Concentration(%)

Fig. 6 Relationship between compressive stress of the sandfixing specimens and concentrations of W-OH
Y=0.17576X - 0.22588 R2=0.97(1)

The objective of the wind tunnel test was to study winderosion resistance. The specimens were made by spraying
W-OH on sand that was contained in a rectangular sand tablewith dimensions of 32 cm×40 cm×4.5 cm. Concentrations
290

Decomposed rate of sand-fixing layer(g)Compressive stress(MPa)

1.2

45

固化砂层的分解重量（g）

3.0

Thickness /cm

24

Density of sand: 1.52 g/cm3
Porosity of sand: 42.6%
thicknessinitial porosity

Decomposed rate of sand-fixing layer(g)3.5of W-OH set by 2%、2.5%、3%、3.5%、4%、4.5%、5%、
6% and 7%. The prepared sand-fixing specimens were placed
into the testing section of the wind tunnel and subjected towind-sand flow under various certain velocities and angles.
The velocity was measured and controlled by the pilot tube,
and a sand sampler with multiple tubes was used to measuresand flow. The testing parameters were set with a wind/sandflow velocity of 10 m/s for 30 min, 15 m/s for 20 min, 20 m/s
for 10 min or 25 m/s for 8 min and angles of 0°, 30° or 60°.
The result demonstrated that W-OH sand-fixing specimens
had excellent wind/sand erosion resistance. The resulting
solidified network can withstand strong wind/sand erosion ata high speed of 25 m/s without deformation or surface
damage. This implies that layers fixed with W-OH canwithstand wind erosion exceeding grade 8 and are completely
suited for use in desert control.

Radiation time (hour)Fig. 7 Relationship between W-OH durability andconcentration of W-US

Weight of evaporation (g)concentration of 0.015% is used and 96% for a W-US
concentration of 0.05%. When the W-US concentration isincreased to 0.1%, the degradation of the sand-fixing layer
with W-OH is almost completely inhibited. Based on the UV
radiation intensity found around Qinghai Lake, China, which
is 600~720 kJ/cm2/a and an experimental UV radiation
intensity of 1000 w/m2, the UV radiation duration of
140.6~166.7 hours in our accelerated aging test is equivalent
to one month in the desert near Qinghai Lake. The service lifeof W-OH without the inclusion of W-US is only one and ahalf months. The service life is increased to over 10 months
with the inclusion of 0.015% W-US, and to 6 years and 8months with the inclusion of 0.05% W-US. When the W-US
concentration is increased to 0.1%, the lifetime increases to
22 years and 8 months, or essentially no degradation in thestrong UV radiation region around Qinghai Lake.
A super water absorbent resin (SAP or W-BW as productname) was added to improve the water retention of the W-OH
sand-fixing layer. It was mixed in the aqueous W-OH
solution and sprayed on the sand surface. The results areshown in Fig.8. Water retention capacity can be improved by
28~35 ％ with the addition of 0.06~0.5 ％ W-BW. By
increasing the concentration of W-BW to 0.7%, the water
retention capacity increases. However, when theconcentration of W-BW is larger than 0.7%, the water
retention capacity increases only slightly. Thus, water
retention capacity can be improved by 50％ with 0.5~1％
W-BW, and 0.7% is the ideal concentration of W-BW for
improved water retention.

a. W-US 0%

b. W-US 0.05%

c. W-US 0.1%
Fig.9 Effect of sand fixation on vegetation.
(After one month growth)
relative germination rate was greater than 80%, indicating afavorable environment for growth. Vegetation growthcontinued to improve with time. Fig. 9 illustrates the growth
conditions after one month. The W-US additive had no effecton vegetation.

Time (min)Fig. 8 Performance of water retention of W-OH
sand-fixing layer

5. Demonstration of sand fixation study with W-OH

4.4 Effect of sand fixation on vegetation
To determine how the sand fixing layer affects vegetation,
Toyoura sand with a density of 1.56 g/cm3 was placed into amold with the dimensions of 30 cm × 30 cm × 5 cm andcompacted by hand. The prepared sand model was 4.5 cm
thick, with a porosity of 40%. A concentration of 3% aqueous
W-OH solution with varying concentrations of W-US was
prepared and homogenously sprayed on the sand surface
homogeneously. The concentrations of W-US used in this test
were 0%, 0.05% and 0.1%, and this experiment wasdeveloped by JCK Co. Ltd, Japan. The seeds germinated one
week after W-OH sand treatment. Two weeks later, the

The W-OH treatment was applied to sand dunes andvegetation in Qinghai China. More than 5 demonstration sites
were built. The effect of sand fixation on vegetation are better
observed in actual application. Fig. 10 shows an experimentconducted near around Qinghai Lake, supported by the
NEDO Program. The demonstrations were performed over 2
years, with constructed areas of 6600 m2 and 33000 m2,
respectively. Weather in this region, which is considered amobile dune, is characterized by high UV radiation, strong
wind and extremely low precipitation. Therefore, traditionalorganic sand-fixing materials readily degrade and are notvery effective. Our experiments demonstrated that W-OH
was a good sand-fixing material, effective in both sand291

verified by field demonstration. In conclusion, the
W-OH-based technology has the potential to be one of themost important and efficient methods for application indesertification control and ecological restoration.
7 ACKNOWLEDGMENTS
This work was financial support by the China-Japan
Cooperation Projects of International Science andTechnology (2010DFB54170). The authors also gratefully
expressed their thanks to Water Conservancy andHydroelectric Scientific Research Institute of Qinghai
Province, for their participating in building demonstrations.

a. Beginning

8 REFERENCES

b. One year later
Fig. 10 Demonstration of vegetation and sand fixation of asand dune around Qinghai Lake, China
fixation and vegetation growth, even under high UV radiationand other adverse conditions. Vegetation growth proceeded
well over the one year treatment and the vegetation coverageincreased from 0 to over 50%. Results were verified by aseparate demonstration in Wudaoliang, along the QinghaiTibet Railway. In this experiment, the vegetation growth was
healthy after a two-year treatment, with vegetation increasingfrom 0 to over 40%. In addition, the W-OH treatment showed
excellent UV radiation resistance [16].
6 CONCLUSION
W-OH is a solidification material with hydrophilic
polyurethane being the fundamental ingredient. W-OH reacts
with water at a broad proportion, resulting in a solidified
network after 2 to 20 min. A 3% W-OH concentration isnecessary for sand fixation. The solid formed does notdissolve in water and boasts large water-holding and water
retention capacity. A sand-fixing layer can be formed byspraying aqueous W-OH solution on the sand. With aspraying amount of 3 L/m2, the resulting sand-fixing layer hasa thickness of 14 mm and a porosity of 20%. The sand-fixing
layer has good elastic properties and mechanical durability,
withstanding wind erosion higher than grade 8. The UV
degradation resistance and water retention can be greatly
improved adding W-US and W-BW to W-OH. Moreover, theUV degradation period can be controlled for a period of 1 to
20 years by adjusting the concentration of the W-US additive.
Therefore, the layer formed provides a good environment forvegetation. The technology for the construction of a W-OH
layer is simple and does not require large mechanicalequipment, making it convenient for large-scale applications.
The effects of sand fixation on vegetation growth were also

[1] UNEP, “United Nations Convention to Combat Desertification,” Nairobi,
Kenya, 1994, 35.
[2] Jian J, Li HJ, Dai XA, “Research on Land Sandy Desertification Using
Remote Sensing Taking Qinghai Lake Area as an Example,”
GEO-INFORMATION SCIENCE, 8(2), 2006, pp.116-120.
[3] Wang LL, Qin ZY, Qin NS, “Climate Change and Its Impact on
Desertification around Qinghai Lake,” PLATEAU METEOROLOGY, 21(1),2002, pp. 59-65.
[4] Dirk G, Zvi YO, “Aeolian dust erosion on different types of hills in a
rocky desert: wind tunnel simulations and field measurements,” Journal of
Arid Environments, 37, 1997, pp. 209-229.
[5] Zhang TH, Zhao HL, Li SG, Li FR, Shirato Y, Ichiro Taniyama TO, “A
comparison of different measures for stabilizing moving sand dunes in the
Horqin Sandy Land of Inner Mongolia, China,” Journal of Arid
Environments, 58, 2004, pp. 203–214.
[6] Li FR, Kang LF, Zhang H, Zhao LY, Shirato Y, Taniyama I, “Changes inintensity of wind erosion at different stages of degradation development ingrasslands of Inner Mongolia, China,” Journal of Arid Environments, 62,2005, pp. 567–585.
[7] Dong ZB, Chen GT, He XD, Han ZW, Wang XM, “Controlling blown
sand along the highway crossing the Taklimakan Desert,” Journal of Arid
Environments, 57, 2004, pp. 329-344.
[8] Ren J, Tao L, Liu XM, “Effect of different microhabitats and stand age on
survival of introduced sand-fixing plants,” Journal of Arid Environments, 51,
2002, pp. 413-421.
[9] FUJIYOSITAKAO, “Integrated areas to prevent the desertification,
Chemistry and Industry,” Chemical Society of Japan, 61(2), 2008, pp.
103-107.
[10] Yang J, Wang F, Fang L, Tan TW, “Synthesis, characterization andapplication of a novel chemical sand fixing agent-poly(aspartic acid) and its
composites,” Environmental Pollution, 49(1), 2007, pp.125-130.
[11] Z. Dong, L. Wang, S. Zhao, “A potential compound for sand fixation
synthesized from the effluent of pulp and paper mills,” Journal of Arid
Environments, 72, 2008, pp. 1388–1393.
[12] Wu ZR, Iwashita K, Wu ZS, Inagaki H, “Research on Sand Dune
Fixation and Green Vegetation of a Novel Chemical Material andConstruction Method of Demonstration Areas Around Qinghai Lake,” The
Report of METI NEDO Project: H19 NO. 0710002, July 2007-Feb.2009,2009[13] Gao WM, Wu ZR, Wu ZS, Yang CQ, Li RJ, Dong ZB, “Study on
Resistance of Wind/sand Erosion Based on a Solidifying Composite Materialof Hydrophilic Polyurethane,” Journal of the Japan Society of Material
Science, 60(9), accepted, 2011[14] Wu ZR, Iwashita K, Wu ZS, Inagaki H, “Experimental study on
evaluation and control of ultraviolet resistance of sand stabilized withorganic slurry containing hydrophilic polyurethane,” Journal of the Japan
Society of Material Science, 57(11), 2008, pp. 1167-1172.
[15] Wu ZR, Wu ZS, Iwashita K, Gao WM, Inagaki H, “UV Irradiation
Degradation Control of Sand Stabilized with Organic Slurry Containing
Hydrophilic Polyurethane,” Journal of the Japan Society of Material Science,60(3), 2011, pp. 235-239.
[16] Wu ZR, “Development of mobile sand-fixing and vegetation techniquewith Hydrophilic Polyurethane,” Ph.D.Thesis, 2011.3.

292

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Migration of Radioactive Materials through Porous Media: Theory and a
Solution for an Unsteady State Problem
Hamza K.I., Civil Engineering Department, Faculty of Engineering, Fayoum University, Fayoum, Egypt



ABSTRACT: Groundwater constitutes a major natural resource
for freshwater supply. Transport of pollutants in porous media issubjected to different processes; they can sorbed onto the surface
on the mineral, sorbed by organic carbon, undergo chemical
reactions …etc. The flow of water is under hydraulic gradient,while transport of solute ions is under concentration gradient. In
this paper, a mathematical development of a two-dimensional
numerical model to simulate density-dependent leachate migrationof radioactive materials in a leaky aquifer system is presented.
Various transport processes such as advection, dispersion,diffusion, sorption and retardation with the concept of reference
density are considered. The model can be used to predict theequiconcentration and equipotentional lines for various cases withdifferent hydraulic and geometric parameters. The influence of thehalf live time of leaching materials on plume migration is
examined, and results show that decreased of pollutant transport
will occur under such conditions.
Keywords: Concentration, Density-dependent, Migration,
Reactive, Reference density, half live time.
1. INTRODUCTION
Groundwater contamination includes any deterioration of thequality of the underground water. Methods of contamination topertinent of the physical environment include wastewaterdisposal practices (both at land surface or deep formulation or
aquifers). Land disposal sites of solid waste can be sources ofgroundwater contamination because of leachate generated bywater percolating through the bodies of refuse, or by refusematerials decomposition.
Many materials dissolve in water, whereas others may be
carried with the water in suspension. Reactive materials
released into groundwater, in connection with nuclear power
productions, by accidents, from landfills, or from nuclear waste
repositories, pose a pollution problem that requires special
attention.
Many studies have been conducted to simulate flow andpollution of groundwater. Reference [16] applied the finite
element theory to approximate the nonlinear equations forvelocity and pressure. The solution for concentration was used
to update the density profile assumed in the flow equation and a
new estimate for velocity and pressure was obtained. Iteration
between the flow and transport equations was preformed untilconvergence was achieved.
Reference [21] employed a two-dimensional hybrid
finite-element method and integrated finite-difference method
to simulate the fluid-density dependent saturated or unsaturated
groundwater flow and transport of solute in porous media(pressure based model). The simulation may be employed forareal and cross-sectional modeling of saturated groundwaterflow systems, and for cross-sectional modeling of unsaturated
flows.
Reference [6] simulated the evolution of the Chloride plume at
an abandoned landfill in a shallow sand aquifer at the
Canandian Force Base Borden, Ontario. A high- resolution grid
was used to obtain an accurate simulation and a consistent flow
field. The simulation was nonunique to some extent because alongitudinal spreading similar to that produced by thelongitudinal dispersivity may also be produced by an increasing
source function.
Reference [9] developed a two-dimensional finite element
model for solute transport in leaky aquifer systems to simulate
pollution migration in heterogeneous anisotropic porous media.
The model was capable of predicting pollutant migration inleaky aquifers under steady and transient flow conditions withpoint and/or line source of pollutions. Reference [7] derived ananalytical solution for the advection-dispersion equation with
rate-limited desorption and first-order decay. The model
equations represent one-dimensional, steady state groundwaterflow through homogeneous isotropic aquifer. It was found thatthe larger desorption and decay rates and the smaller the
distribution coefficient, the faster the rate of contaminantremoval from the aqueous and sorbed phases.
Reference [17] presented a two-dimensional cross sectional
finite difference model which can simulate density dependent
pollutant migration from dumped disposal sites in multilayer
leaky aquifers using the concept of reference density. They
found that high equiconcentration lines did not movesignificantly after 2000 days, while low equiconcentration lines
continued to advance with time.
Reference [18] developed a numerical model to study the mixed
convection processes below a saline disposal basin locatedbetween a recharge and discharge zone. Numerical simulations
are performed in cross-section using the 2-D density dependent
model SUTRA [21]. Both homogenous and heterogeneous
aquifer systems were studied and the effects of anisotropy wereconsidered. The movement of salt to the underlying
groundwater system from Lake Tutchwwop, a saline disposal
complex in north-central Victoria was modeled in cross-section[19]. Density-dependent flow was simulated due to salinity
contrast between the hypersaline bain waters and the regional
groundwater.
Reference [10] introduces a two-dimensional cross section
finite difference model, which can simulate scale dependent
dispersion for the transport of the dissolved substances ingroundwater aquifers. The dispersivity coefficient was taken to293be dependent on the travel distance of the solute from an input
source.
A series of column experiments was undertaken to explore theinfluence of colloid input concentration (2, 1, 0.5, and 0.25
times a reference concentration, colloid size (negtivellychanged 3.2 and 1.0 m quartz sands) on transport anddeposition [5]. They found that for a given input concentration,
decreasing the sand size and increasing the colloid size resulted
in increased mass retention in the sand near the column inlet andlower relative concentrations in the effluent. A finite-element
code COMSOL was used by [8] to model dispersion in
two-dimensional artificial packed beds. The steady state
incompressible Navier-stokes equation along with theconvection diffusion equation was solved to obtain aninteresting analysis of the effect of porosity and heterogeneity
on dispersion coefficient.
In this paper, the finite element model of incompressible fluid
flow in porous media developed by [11] was extended to model
dispersive of reactive materials in a leaky aquifer system. The
flow and transport equations are coupled, using the reference
density approach with the elimination of the intermediate stepof calculating velocities. The developed model is a
two-dimensional fluid density-dependent finite element model
for reactive materials (2D-FERM). Leachate migration ofreactive materials in leaky aquifer system is simulated.
Governing equations, including flow and transport equations,are coupled using a new approach with replacing fluids withdifferent densities by hypothetical fluid (reference density
approach). The dispersion coefficients are considered to bevelocity dependent, and their components values are updated ineach iteration. The (2D-FERM) model is then used to predict
the shape of equiconcentration and equipotential lines in leaky
aquifer system.
2. VARIABLE CONTAMINATION PATTERNSThe complex variable to be considered in evaluating thepotential of contamination to spread in groundwater calls forselected concepts. A usual concept centers on two opposing
tendencies of the movement of contaminants in the ground. One
is the tendency of the contaminant to move with groundwater,
and the other is for contaminates to be retarded or degraded bymechanisms of attenuation. Different contaminants attenuate tovarying degrees by: (1) dilution with water; (2) decay with timeor some other degradation mechanism, and (3) sorption on theearth materials [1]. For example, chlorides may attenuate only
by dilution, but certain radioactive nuclides attenuate
effectively by a combination of dilution, decay with time, andsorption on clays [2].
Questions center on the extent to which one can rely on sorption,
dilution, and/or decay to allow contaminants to disappear or
reach a harmless levels some distance from points of water use.
In some cases contaminates do not extended more than a few
meters from waste sites, whereas in other cases, they may
extend for long distances in the direction of groundwater flow.

3. REFERENCE DENSITY APPROACH
The density of water in natural aquifer systems ranges from
998.2 Kg/m3 at 20oC for freshwater to greater than 1,345
Kg/m3 as reported for the salado brine of New Mexico.
Reference [12] set a classification for saline water, ranging
from slightly saline to brine. He divided the salty water into
four groups on the basis of total dissolved solids as slightly
saline, moderately saline, very saline, and brine.
The density of the flowing groundwater varies according to its
concentration. As such, piezometric heads at various nodes
would not indicate the actual pressures unless associated withindigenous water density. On the other hand, ignoring theeffect of density variation may result into false or misleading
conclusions, with specific reference to sensitive problems
where small variations in fluid density affect the flow andtransport regimes considerably. Leaching from dumped
disposal sites is a good example for such cases. In suchsituations, it is more convenient to employ the concept ofreference density. Various densities are unified to the
freshwater density and equivalent freshwater heads areevaluated. Finally, one reference density for the whole domain
is defined and all the piezometric heads are evaluated interms of freshwater heads.
4. GOVERNING EQUATIONS
The governing equations describing the radioactive materials
transport in groundwater under unsteady state conditions are asin [4]:
1. The general Darcy equation for groundwater flow:k(1)
q = - (P + g Z)where,
q is the specific discharge vector (LT-1),
k is the intrinsic permeability tensor (L2),μ is the dynamic viscosity (ML-1T-1),
P is the pressure (M L-1T-2),
ρf is the fresh water (reference) density (ML-3),
ρ is the fluid density (ML-3),
g is the gravitational acceleration (LT-2),
Z is a space coordinates (L).
kf g

p
Substitution of h =+ z, K =
, and
-1
r=
g

f
f
into "(1)," yields

q = - K (  h +  r Z)(2)

where,h is the equivalent hydraulic head (L),K is the hydraulic conductivity tensor (LT-1),
ρr is the relative density (L0).
2. The mass balance equation for the fluid can be written as:

294

n
= - . q + R  ret(3)

where,n is the effective porosity (Lo),t is the time (t),R is a recharge rate per unit volume of aquifer medium (T-1),
ρre is the density of the recharge water (ML-3).
3. The hydrodynamic dispersion equation or the mass balance
equation for the pollutant can be written as:
CnC= - . (Cq - nDh .C) + (1 - n) K d  btt- (1 - n)  b K s F - n C  RC r(4)where,C is the pollutant concentration (ML-3),
nC is the quantity of the pollutant increased within a control
box (ML-3),
Dh is the hydrodynamic dispersion coefficient (L2T-1) which
can be calculated as in [3]:D xx =  1

V2
V 2x+  t z + D*d ,
|V |
|V |

D zz =  t

V 2X
V2+  1 z + D*d , and
|V |
|V |

V xV zD xz = D zx = (  1 -  t )V(5)

where,
α1 is a longitudinal dispersivity (L),is a transversal dispersivity (L),αt
Vx is pore-velocity in x direction (LT-1),
Vz is pore-velocity in z direction (LT-1),
V is magnitude of the resultant velocity (LT-1),
Dd* is the coefficient of molecular diffusion (L2T-1),
Cre is the recharged water concentration (ML-3),
Kd is the distribution coefficient,ρbis the bulk density of the porous medium (ML-3),
Ks is the degradation rate constant of the water (T-1),
λis equal to 1/T, where T is the half live time (T).
F
is the mass of pollutant on the solid per unit mass of bulk
dry porous medium which is defined as F = Kd C.
4. A constitutive equation relating fluid density to contaminantconcentration can be expressed as: =  o  a(C  Co )(6)

where,
Co is the reference concentration (ML-3),
a is a known constant (Lo) called coefficient of fluid densitychange which can be calculated as: max -  o(7)a=
C max - C owhere, ρmax is the maximum water density corresponding to
the maximum concentration Cmax.

5. BOUNDARY CONDITIONSConsider the idealized leaky aquifer domain as shown in Fig.
1. A line source of pollutant, represent the dumped disposal
site, is placed at the bottom of the aquitard. The aquifer isrecharged by freshwater entering at the left boundary B4 andby downward leakage through the upper boundary B1, where
the water table is higher than the aquifer piezometric head.
Water, entering into the system, flows out of the systemthrough the right boundary B2.
The downward flux through upper boundary B1 is governedby Darcy’s equation. The concentration below the pollutant
source will be set equal to the source concentration, Cmax.
Solute particles are assumed to move only vertically throughthe aquitard. Along the remaining part of boundary, theconcentration is set equal to freshwater concentration. The
right boundary B2 allows for advective transport of pollutant
ions only, i.e., the concentration gradient is set equal to zero.
This boundary is subject to hydrostatic pressure distribution.
The pressure distribution along B2 is not constant but varies as
contaminant concentration increases when the leachate plume
approaches boundary. The piezometric head at this boundary istransformed into equivalent freshwater head, and its value isupdated as the simulation proceeds according to the newconcentrations at this boundary. The bottom boundary B3 isimpermeable, i.e., the normal flux through it for both fluid andpollutant ions are equal to zero. The left boundary B4 islocated where the concentration is assumed to be constant andequal to freshwater concentration, Cf. The pressure along thisboundary is hydrostatic. Therefore, the boundary conditions
can be written as follow:
 On the top boundary B1
 W.T.  P.H.
 r 
q z  K z\ bC = Cmax below the pollution source,otherwise
C = Cf\where, K z is the vertical hydraulic conductivity for the aquitard(m/s), b is the aquitard thickness (m), W.T. is free water table
elevation (m), P.H. is the piezometric head throughout theaquifer, and r is the relative density.

 On the right boundary B2
h = hsC0xwhere, hs is the piezometric head at the right boundary.
 On the bottom boundary B3
qz = 0C0z On the left boundary B4
h = hoC = Cfwhere, ho is the piezometric head at the left boundary.

295

7. MODEL APPLICATIONS

Ground surface
Water tablei= 1 .5 * 1 0-2Peizometric head
Aquitard
C=CmaxB1
C=0.5 kg/m3

depth =40m

C=0.5 kg/m3

Kxx = 0.07 m/min
C=0.5 kg/m3

L = 10m

h=ho

C
0
z

B4

C
0
x

K zz = 0.014 m/min

T = 1.0m

Impermeable B3

B2

C
0
z

h=hs

Bottom Boundary (q=0)
Aquifer length, 200m

Figure 1: Idealized leaky aquifer system and boundary
conditions

6. MATHEMATICAL DERIVATION OF THE(2D-FERM) MODEL
Because the contaminant transport problem is a nonlinear one,and the velocity field and hence the dispersion coefficientsdepend upon concentration distribution, Galerkin's procedure iswell suited for the problem [15]. The Galerkin finite element
technique is therefore adopted. Triangular elements are usedwith linear interpolation functions, <N> = <Ni Nj Nk>.
Application of the Galerkin technique with Gauss's theorem tothe fluid mass balance equation in two-dimensions yields:
 < N >T 1
 < N >T 1
[ (  q z )] dAe[ (  q x )] dAe +  Ae
 Ae
X
z
nq
nq
(
)

q
C
n
+  Ae< N >T
=  Le < N >T
(8)
dLe
nq
t

To determine the shape of equiconcentration and equipotential
lines in a leaky aquifer, the model is applied to the system
shown in Fig.1. A study domain, 200x40m in area and overlain
by a semi-pervious layer 5m thick, was chosen. The aquifer
hydraulic conductivities in the x and z directions, Kxx and Kzz,
were taken as 0.07 m/min and 0.014 m/min, respectively. The
vertical hydraulic conductivity of the upper semi-pervious layer,
Kz\, was set to equal 2.8x10-3 m/min. The slope of the hydraulicgradient is assumed to be 1.5x10-2. Reference [20] indicated thatvalues of longitudinal dispersivity, αL, which have beenreported in field investigations range from 6.0m to 150.0m.
Lateral dispersivity, αT, has usually been taken to be 10-30% of
the longitudinal values [14]. Owing to the scale behavior ofdispersion, no generally valid values for longitudinal
dispersivities can be given [13]. Values between 0.1 and
500.0m can be found in the literature. The Longitudinal andtransversal dispersivities, αL and αT, are taken as 1.0m and 0.1m,
respectively. The effect of molecular diffusion was neglected.
Model parameters are given in Table 1.
Table 1: Model parameters for the study caseFreshwater density,  =1000 kg/m3
Groundwater concentration, Cf = 5000 mg/l
Contaminate concentration, Cmax = 500000 mg/l
Dynamic viscosity,  =0.001 kg/(m.s)
Maximum density of recharge water, max =1035 kg/m3
Coefficient of fluid density change, a = 0.707
Water compressibility = 4.5x10-10 Pa-1
Soil compressibility = 1.0x10-8 Pa-1
Porosity, n = 0.3
Hydraulic conductivity, Kxx= 0.07 m/min
Hydraulic conductivity, Kzz= 0.014 m/min
Longitudinal dispersivity,  L = 10.0m

Similarly, the two dimensional hydrodynamic dispersion
equation can be written as:

< N > 1
CC[n( D xx
+ D xz) - q x C] dAe
X
x
z
nR
T
1

C
C
+  Ae < N >[n( D xz
+ D zz) - q z C] dAe
nR
z
x
z
(1 - n)  b K s K d C enC e-  Ae < N >T
dA -  Ae < N >TdAnRnR
< N >T
C e(9)+  Ae< N >T(n D h C - qC)n dL e
dA =  LenR
t
Equations (8) and (9) can be written symbolically in the matrix
form as:(10)[Fh ] {h k 1  {A} Ae

Transversal dispersivity,  T = 1.0m
Acceleration due to gravity, g = 9.81 m/s2
Molecular diffusivity, Do = 0.0

T[FC ] {C k 1  {B}(11)

Where [Fh] and [FC] are the coefficient matrices, {hk+1} and{Ck+1} are the unknown hydraulic heads and concentrations at
the new time level k+1, k is the time level, and {A} and {B} are
the known vectors which contain the effect of boundaries.
Equations (10) and (11) constitute the 2D-FERM model.

A contaminant source is provided at the bottom of the aquitard
in the range of 16<x<48m. Below the pollution source, theconcentration was set equal to the maximum concentration,
Cmax. Otherwise, the concentration was set equal to the
freshwater concentration; Cf. The effective porosity of 0.3 was
employed throughout the aquifer. In the aquitard it was set toequal 0.4. A Dirichet type boundary condition was used at the
top and the left boundaries, while a Neumann type boundary
condition was assumed at the remaining boundaries. The
application of Neumann type boundary condition for transport
at an exit boundary is acceptable only when the plume meets theboundary at right angle; this condition is satisfied here, as theexit boundary is located away from the contamination source.
Flow, through the upper semi-confining layer (aquitard) was
simulated by the one-dimensional form of Darcy's equation.
The hydraulic gradient through the aquitard was calculated by296

dividing the difference between the water table above theaquitard and the piezometric head below it by its thickness. The
study domain was represented by finite-element grid has 390
nodes, defining 696 triangular elements, on a rectangularspacing. The time step was constant at 20 days. Figure 1
presents the study domain and hydraulic parameters of this run .
The convergence criterion, ε, was set equal to 1x10-3. The
matrix solver does not need iterative procedures because it issolved by a modified Gauss elimination method, andincorporating the banded matrix to reduced the storage space.
The simulation was carried out after 5 and 10 years ofcontinuous leaching. Fig. 2-a and Fig. 2-b present thecorresponding equiconcentration and equipotential lines,
respectively. Equiconcentration line 0.3 of the maximum
concentration (50.0 kg/m3) advanced 85m measured from theleft boundary after 5 years of leaching while equiconcentrationline 0.1 advanced 119m, and reached a maximum depth ofabout 22m after the same period of time as shown in Fig. 2-a.
After 10 years of continuous leaching as shown in Fig. 2-b,
equiconcentration line 0.3 advanced 116m from the left
boundary, while equiconcentration line 0.1 traveled a distanceof 176m and reached a maximum depth of about 30m.
The freshwater recharge from the left and top boundaries mixes
with denser polluted water before discharging from the rightboundary. The bottom boundary represents no-flow boundary.
The aquifer, initially, was assumed to contain freshwater,
therefore contour lines of equivalent freshwater head werevertical and orthogonal to the top and bottom boundaries withequal distances for the equal head intervals, as shown near theright boundary (Fig. 2-a).
Transient and steady states distributions of equivalent
freshwater heads were obtained by simulating the coupled flow
and solute transport using the developed model. Contours ofequivalent freshwater heads were distorted, as shown in Figs.
2-a, and b, except in the right part of the domain where its
pollutant concentration was relatively low. In the upper left side,
where the pollutant source was placed and the fluid densities
were high, values of the equivalent freshwater head was alsohigh. It was noticed that the effect of pollutant concentration on
the fluid density was considerable and increased with time.
Reactive materials are defined with their half live times (time atwhich the materials loss 50% of their concentration). The
developed model was used to simulate behavior of thesematerials with a half live time of 2000 days. The hydraulic andgeometric parameters were kept the same as in the previous
case. Fig. 3-a, and Fig. 3-b present the shape of
equiconcentration lines after 5 and 10 years, respectively.
Equiconcentration line 0.1 advanced 67m, and reached amaximum depth of about 10.5m after 5 years of leaching, asshown in Fig. 3-a. After 10 years, as shown in Fig. 3-b,
equiconcentration line 0.3 advanced 68m beyond from the left
boundary, while equiconcentration line 0.1 advanced about
89m and reached a maximum depth of about 15.5m.
It was found that higher equiconcentration lines (0.7 and 0.9 ofmaximum concentration) reach the steady state condition earlier

than the low equiconcentration one (0.1 and 0.3 relative
concentration). In case of reactive materials the plume occupied
smaller area than the non-reactive materials.

40

0

20

40

60

80

100

120

140

160

180

200
40

30

30

20

20

10

10

0

0

20

40

60

80

100

120

140

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200

40

60

80

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140

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180

200
40

a- after 5 years
40

0

20

30

30

20

20

10

10

0

0

20

40

60

80

100

120

140

160

180

0
200

b- after 10 years

Figure 2: Equiconcentration and equipotential lines (without
radioactive materials)

40

0

20

40

60

80

100

120

140

160

180

200
40

30

30

20

20

10

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180

0
200

40

60

80

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200
40

a- after 5 years
40

0

20

30

30

20

20

10

10

0

0

20

40

60

80

100

120

140

160

180

0
200

b- after 10 years

Figure 3: Equiconcentration and equipotential lines (withradioactive materials)

8. CONCLUSIONS
Transport phenomenon in groundwater involves several
processes which are driven by density variation, convection
and dispersion. The flow of water in the aquifer is underhydraulic gradient, while the transport of solute ions is underconcentration gradient. The dependency of the dispersion
coefficients, in the hydrodynamic dispersion equation, on thevelocity of fluid and concentration distribution makes the
problem nonlinear, and hence requires numerical solutions.
Mathematical development of a two-dimensional finite-element
model to simulate density dependent leachate migration ofreactive materials in leaky aquifers is presented. The concept ofreference density, which employed in the model, increased
accuracy (over pressure based models) in representation of thetransport process. The flow and transport equations are coupledwith the elimination of the intermediate step of calculating
velocities. The numerical scheme is stable and convergent.

297

The applied model is stable and the proposed numerical scheme
converges under different flow and boundary conditions. It was
found that low equiconcentration lines (0.1 and 0.3) continued
to move for a long period of time after the pollution source wasplaced, while higher equiconcentration lines attained the steadystate after about earlier than the low equiconcentration lines inmost of the cases. Overlapping the equiconcentration lines on
the equipotential lines indicated that high concentrations wereassociated with higher potential heads. The half live time forleachting materials is the most important parameter such that it
predominates the migration process.

[20] Volker, R.E. and Rushton, K.R., “An assessment of the importance ofsome parameters for seawater intrusion in aquifers and a comparisonof dispersive and sharp-interface modeling approaches”, Journal of
Hydrology, 56: 239-250, 1982.
[21] Voss, C.I., “SUTRA- A Finite Element Simulation Model for
Saturated-Unsaturated Fluid Density Dependent Groundwater Flowwith Energy Transport”, U.S. Geol. Surv., Water Resources. Invest.
Rep., 84-4262, 406pp, 1984.

9. REFERENCES[1]

Anderson, J. R., and Dornbush, J. N., “Influence of sanitary on
groundwater quality”, Journal of American Water Works Association
AWWA, Vol. 59, April 1967.
[2] Bagchi, A., “Design of natural alternation landfill”, Journal ofEnvironmental Engineering Division, ASCE, Vol. 109, No. 4, Aug.,1983.
[3] Bear, J., “Hydraulics of Groundwater”, New York, McGraw-HillBook Co., 1979.
[4] Bear, J., and Verruijt, A., “Modeling Groundwater Flow and
Pollution”, Redial Publishing Company, Dordrecht, Holland, 1987.
[5] Bradford, S.A. and Bettahar, M, “Concentration Dependent Transportof Colloids in Saturated Porous Media”, Journal of Hydrology, 82:
Issue 1-2, 99-117, 2006.
[6] Frind, E.O. and Hokkanen, G.E., “Simulation of the Bordon plumeusing the alternating direction Galerkin technique”, Water Resources
Research, 23(5): 918-930, 1987.
[7] Fry, V.A., Istok, J.D., and Guenther, R.B., “An analytical solution to
the solute transport equation with rate-limited desorption and decay”,
Water Resources Research 29(9) 3201-3208, 1993.
[8] Garmeh, G. Johns, R.T., and Lake, L.W., “Pore-scale Simulation of
Dispersion in porous Media”, SPE J. Dec., 559-567, 2009.
[9] Hamza, K.I., “A two-dimensional finite element model for solutetransport in groundwater”, M. Sc. thesis, Faculty of Engineering,
Cairo University, Egypt, 1993.
[10] Hamza, K.I., “Scale-dependent dispersion model in porous media”,
Proceedings of the International Conference on Geoenvironment 2000
(ICG200), Sultan Qaboos University, Muscat, Sultanate of Oman,
March, 4-7, 2000.
[11] Hamza, K.I. “A Two-dimensional finite element model for pollutant
transport out of dumped disposal sites: (2D-FEPT)”, Dirasat, anInternational Journal for Engineering Science, University of Jordan,
Jordan, 31(1): 82-96, April 2004.
[12] Krieger, R. A., Hatchett, J.L. and Poole, J.F., “Preliminary Survey
ofthe Saline-Water Resources of the United States”, U.S. Geological
Survey, Water Supply Paper no. 1374, 1957, 172 pp.
[13] Kinzelbach, W., “Groundwater Modeling”. Elsevier, New York, 333
PP, 1986.
[14] Pickens, F.J. and Grisak, E.G., “Scale-dependent dispersion in
stratified granular aquifers”, Water Resources Research, 17 (4): 1191121, 1980.
[15] Pinder, G.F., “A Galerkin finite-element simulation of groundwatercontamination on long island”, Water Resources Research, 9(6): 16571669, 1973.
[16] Segol, G., Pinder, G.F., and Gray, W.G., “A Galerkin finite element
technique for calculating the transient position of the saltwater front”,
Water Resources. Research, 11(2): 343-347, 1975.
[17] Sherif, M.M., and Singh, V.P., “Leachate migration in multilayer
aquifers”, Proceedings of the International Conference, “Transport andReactive Processes in Aquifers”, Rotterdam, Netherlands, 1994.
[18] Simmons, C.T. and Narayan, K.A., “Mixed convection processes
below a saline disposal basin”, Journal of Hydrology, 194: 263-285,
1997.
[19] Simmons, C.T. and Narayan, K.A., “Modeling density-dependent flow
and solute transport at the Lake Tutchewop saline disposal complex,
Victoria”, Journal of Hydrology, 206: 219-236, 1998.

298

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Fundamental Study on Ecosystem Support Canal using Porous Concrete
Masuma Yoshihiro, Graduate School of Integrated Arts and Sciences, Kochi University
Hasegawa Yuki, Graduate School of Integrated Arts and Sciences, Kochi University
MD.ROKUNUZZAMAN, Graduate School of Integrated Arts and Sciences, Kochi University
Yamada Toshio, Water Environment Section, Kaihatsu Concrete Co., Ltd.
Kassai Hirofumi, Water Environment Section, Kaihatsu Concrete Co., Ltd.
Sato Shushi, Faculty of Agriculture, Kochi University



ABSTRACT: This research aimed to enhance the compressive
strength of porous concrete as well as to develop the porous concretethat can support and improve the ecosystem preservation by itself.
Several porous concrete specimens were prepared for themeasurement of mechanical properties. As a result, it was confirmedthat the radius of coarse aggregate affected significantly tomechanical properties of porous concrete under the same unit weightof cement. It was also revealed that strengths at age 28 days werestable despite of different sizes of coarse aggregate. The
bio-adhesive ability of porous concrete specimen was evaluatedagainst water bugs and adhesive algae. Every porous concrete
specimen was soaked in same environmental condition at the bottomof actual concrete canal. From this experiment, it was confirmed that
preference environment for some specific species of water bugs are
possible to be supplied when the porosity and the size of coarse
aggregate would be adjusted.

Keywords: Porous Concrete, Ecosystem Support,
Biodiversity, Bio-adhesive

hydraulic structures such as canal and head works due to their
low permeability and structural safety. On the other hand,
there are a lot of studies concerns to porous concrete thatgives newly function such as environmental consideration.
Up to now, porous concrete becomes to be famous material
for revetment of river bank in order to support the recovery ofvegetation at the river embankment. However, this material
has not used frequently because of its low strength anddurability.
This research aimed to expand the utilization of porous
concrete that has both high strength as well as the special
function that is called as “bio-adhesive ability”. The former
point, to improve the compressive strength of porous concreteas high as structural material, was examined from theviewpoint of strengthening. The latter point, to observe therelationship between the different types of porous concrete
and inhabited creatures, was examined from the viewpoint ofcolonized living things.

1. INTRODUCTION
In Japan, previous improvement programs on agricultural
facilities were mostly executed focused to the functions ofwater use and structural safety, and they were not consideredenough against nature and ecosystem. As a result, various
ecosystems were lost or damaged seriously, it is necessary to
make actions for conservation and recovery of ecosystem
now a day.
From such background, Land Improvement Act wasrevised in 2001 in Japan. In the law above, “consideration to
make the harmony with environment” was provided as theprinciple of every project. However, existing technology thathas such consideration against environment is not establishedcompletely including the evaluation of its effect yet. It isnecessary to analysis from the viewpoint of general judgment
containing the quantitative evaluation on the influence toecosystem.
Concrete materials are mainly used for the agricultural

2. METHOD AND PROCEDURE
2.1 Evaluation on Compressive Strength of Porous
Concrete
Cylindrical porous concrete specimens were placed using
different grain size of coarse aggregate; No.5 (13 – 20mm),
No.6 (5 – 13mm) and No.7 (2.5 – 5mm). Fundamental
mixture proportions of porous concrete are shown in Table I.
Amount of mortar was adjusted to be constant in order toexamine the influence of grain size of coarse aggregate tocompressive strength. Using the biaxial forcing mixer,
concrete was mixed under dry condition for 1 minute andadditional 3 minutes after adding the water. Water amount
was finely adjusted with confirming the adhered mortar
condition around the aggregate particles. Porosity of porous
concrete were varied with changing the packing amount into
mold, and 6 types of porous concrete were placed in thisexperiment. Three numbers of cylindrical specimens for eachtype were tested for measuring the compressive strength.
Handy vibrator was used for the compaction. Standard curing,299

2.2 Summary on Bio-adhesive Characteristic
Experiment of Porous Concrete
Porous concrete specimens sizing of 230mm length 230mm width  60mm height for bio-adhesive
characteristic experiment were made by same processdescribed at previous section. Details of the specimens areshown in Table II. All the specimens were set in actual open
channel for agricultural placed in Kochi University, and werestudied for the evaluation of bio-adhesive ability such as
number of water bugs and amount of attached algae. Concrete
specimens were mentioned to be sunk in the water all the time.
Water bugs settled in porous concrete specimen werecollected on the tray fulfilled with water. Gathered water bugs
were investigated as follows; identify the species using
technical book of water creatures [1]-[4], divided into eachspecies and counted their number. However, bugs less than
1mm of body length were excluded to count.
Algae were gathered with shaving from one surface(downstream side) of specimen. Suspensions were collected,
filtrated with glass filter and measured the wet weight, dryweight and chlorophyll a (call “chl.a” here after). Dry weight
was measured after the treatment of drying 24 hours with
60℃. Absorptiometer was utilized for the measurement ofchl.a.
3 RESULTS AND DISCUSSION
3.1 Compressive Strength of Porous Concrete
Relation between compressive strength and porosity was
shown in Fig.1. Under the same porosity condition,
compressive strength was higher as the larger particle
diameter of coarse aggregate. It was considered that specificsurface of coarse aggregate was decreased when using thelarger sizes of coarse aggregate, and the thickness of mortar
binder surrounding the coarse aggregate particle in unit
volume became increased. Moreover, relationship between
compressive strength and porosity made with coarse
aggregate No.5 and No.6 showed similar value and tendency
despite the thickness of mortar between coarse aggregate
No.5 and No.6 were different. It was revealed that developed

Table I Fundamental Mixture Proportions of Porous Concrete
Grain size
Particle diameter (mm)No.5
13 - 20

No.6
5 - 13No.7
2.5 - 5

Cement (kg/m3)

221

221

221

Admixture (kg/m3)
Sand (kg/m3)
Gravel (kg/m3)

39
233

39
233

39
233

1548

1540

1537

Water (kg/m3)
Porous diameter (mm)57
2.0 - 14.6

57
0.8 - 9.5

57
0.4 - 3.7

Table II Porous Concrete S pecimens detail
Crashed stone size
Particle diameter (mm)At (%)
22.3
No.5
13~20
29.4
22.2
No.6
5~13
25.7
No.7

2.5~5

18.9
20.5
22.7

compressive strength would be stable and enough when theCompressive Strength(N/mm2)

sunk in 20 ℃ water, was applied and measured thecompressive strength at age 14 and 28 days.
Here describes the measurement method on porosity ofporous concrete as follows. Firstly, volume of specimen (V1)
was measured. The specimen was soaked into water for 24hours in order to absorb the water, measure the weight inwater (W1). Secondly, the specimen was put out toatmosphere where the temperature and relative humidity
were controlled into constant condition of 20℃ and 60%,
respectively. Wait until the weight into stable. Furthermore,
set the same specimen for 24 hours in atmosphere under thesame temperature and relative humidity condition wasmeasured as W2. Total porosity was calculated using (1) isshown below;
Total porosity At (%) = (1-((W2-W1) /ρ)/V1)  100(1)
where ρis the density of water.

No.5-14days

No.5-28days

No.6-14days

No.6-28days

No.7-14days

No.7-28days

45
40
35
30
25
20
15
10
10

20
Total Porosity(%)

30

Fig.1 Relation of Compressive Strength and Porosity

thickness of mortar was insured to be standardized.
On the contrary, compressive strength was declined as theincrement of porosity of porous concrete. The main cause ofthis result was occurred due to the reduction of cement
amount in unit weight. Comparing the strength developmentof age from 14 to 28 days, each increment of compressive
strength were evaluated as 7.9% in No.5, 10.0% in No.6, and2.5% in No.7. From this result, it was clarified that strength
development of No.7 was smaller than those of No.5 andNo.6. It meant that strength of No.7 was close to stablecondition while it was age 14 days.
All the porous concrete could gain higher than 20 N/mm2,
as high and stable as normal concrete. It meant that porous
concrete could be useful for the member of structure if therelationship between compressive strength and porosity wasclarified enough.

300

3.2 Bio-adhesive Characteristic Experiment of Porous
Concrete
3.2.1 Species Number and Population Size of Water Bugs
Generic names and species gathered from porous concrete
specimen were shown in Table III. Twenty three of genericnames and twenty six of species were confirmed, and thedominant species was Asellus hilgendorfi. Second, third andfourth dominants were Ectopria, Stenelmis andEphemeroptera, respectively. The change with time of
population size gathered from the specimen was shown in
Fig.2 a). The larger population size was confirmed when sizeof coarse aggregate was bigger. It was considered that various
lengths of body of water bugs were easy to enter to the void of
porous concrete surface when the diameter of porosity waslarge enough. On the other hand, there was a sharp drop in
population size of water bugs during May 21st of 2010 to July
18th of 2010. Several times of torrential rainfall weregenerated before July 18th of 2010, the water bugs were
estimated to be washed away or evaded due to violent waterflow and muddy water.
Change with time of species number in each specimen wasshown in Fig. 2 b). Lots of species were confirmed andvariation of population sizes was larger at the beginning of
the experiment. However, species number was getting stable
as the time went on. It meant that water bugs suited to theenvironment established with porous concrete were selectedwith the time passage, and particular species were becoming
to be the dominant. Judging from the investigation result in2011 and 2010, different species were confirmed. Then it wassuggested that observed species in the past data would be
happened to reach to the specimen from upstream. Further
observation will be necessary whether existing species as adominant were climax species or not. Much population of
Ephemeroptera and Ptilodactylidae were confirmed in theNo.5-22.3
No.6-25.7
No.7-22.7

Table III S pecies List of Water Bugs
Order

Scientific Name
Baetis sp.

Ephacerella longicaudata
Uracanthella punctisetae
Caenis sp.
Leptocerus sp.

a) Population Size
2500

Population Size

Ephemeroptera

Ju.

Sep.

500

Molanna moesta

0

Ecnomidae sp.

10b) Species Number
8

Species Number

Goera japonica
Nemoura sp.
Conchapelopia japonica
Psychodidae sp.
Chironomus samoensis

Hemiptera

Isopoda420Aphelocheirus nawae

1.0c) Diversity Index

Diversity Index

Ectopria opaca
stenelmis sp.
stenelmis sp.

Odonata

6

Aphelocheirus vittatus
Ectopria opaca

Coleoptera

Ju.

1000

Goerodes sp.

Diptera

Mar.

1500

Nippoberaea gracilis

Plecoptera

Dec.

2000

Hydroptila sp.

Trichoptera

No.6-22.2
No.7-20.5

3000

Ecdyonurus sp.
Cinygmula sp.

No.5-29.4
No.7-18.9

Calopterygidae sp.
Gomphidae sp.
Asellus hilgendorfi

0.8
0.6
0.4
0.2
0.0
3/19 5/21 7/18

1/14

4/24

Fig.2 Change with Time of Result
for Water Bugs

301

7/8

class4 CONCLUSION
The results in this study were concluded as follows;
1) Larger particle lead higher compressive strength wheneach porous concrete was used same amount of cement.
2) Higher porosity of porous concrete generated its
compressive strength to be lower when each porous concrete
was used same amount of cement.
302

Table IV S pecies List of Algae
Scientific Name
Coscinodiscus sp.
Melosira sp.
Cyclotella meneghiniana
Hysrosera whampoensis
synedra sp.

Bacillariophyceae

Chlorophyceae

Fragilaria sp.
Naviculaceae sp.
Cymbella tumida
Gomphonema sphaerophorum
Rhoicosphenia abbreviata
Surirella sp.
Nitzschia trybionella
Bacillaria paradoxa
Cocconeis placentula
Cladophora sp.

Quantity of Chl.a (mg chl.a)No.5-22.3
No.6-25.7
No.7-22.7

No.5-29.4
No.7-18.9

No.6-22.2
No.7-20.5

0.8a) Quantity of Chl.a
0.6

Ju.

Sep.

Dec.

Mar

Ju.

0.4

0.2

0.0
0.3

Dry weight (g)b) Dry weight
0.2

0.1

0.0
10

c) Ratio
8

Ratio

porous concrete of No.5 and No.7, respectively. Moreover,
specimen of lower porosity showed highly number of eachspecies. Therefore, it was suggested that coarse aggregate of
No.5 and No.7 might create the preference environment for
Ephemeroptera and Ptilodactylidae, respectively.
Diversity index for each types of specimen was calculated.
Simposon’s formula shown at (2) was chosen to get thediversity index. This formula aims to see the balance between
the species number and population size of biotic community,
shown as follows;(2)D
where Pi is proportion of individual number of species (i) toall individual number of communities, and S is the species
number of objective community. If one species was
extremely superior to other species, diversity index would bevery small. In this study, the population size of Asellus
hilgendorfi was too much, therefore, this dominant was
excluded from the calculation of this index. Change with timeof diversity index was shown in Fig. 2 c). Diversity indexes
were varied in all specimens during the initial period of thisexperiment. However, they came to be stable and similarvalue in all specimens since May of 2011. We confirmed in2011 that some newly species were colonized that madepermanent resident. That could be one reason of stabilizationof that index described above.
3.2.2 Bio-adhesion of Algae Grew on the Porous Concrete
Adhesion of algae was confirmed in all porous concrete
specimens, the name of species was shown in Table IV.
Identified algae were mostly the same kind of diatom(Bacillariophyceae), maximum species number was obtained
at 22.7% porosity of porous concrete using coarse aggregate
No.7. Change with time of quantity variation of chl.a and dryweight was shown in Fig. 3 a) and b), respectively. Quantityof chl.a and dry weight showed same tendency with theresults of water bugs, these values came to be stable as time
passing by. There was no relationship between the quantity ofchl.a and dry weight; calculated correlation coefficient wasalmost zero. Here showed the result of relationship between
time and ratio (quantity of chl.a / dry weight) in Fig. 3 c).
Assuming that quantity of chl.a was an indicator of living
algae as well as dry weight was that of both living and deadalgae, variation of this ratio was very large in most of all
specimens. However, in 18.9% porosity of porous concrete
specimen using coarse aggregate No.7, this ratio was stablecomparatively. It was suggested that this type of porous
concrete would be good adhesion ability against diatom. The
reason was considered that smaller particles of aggregate
gave larger specific surface and possible to give better
condition for the growth of this algae.

(Quantity of Chl.a/Dry weight)

6
4
2
0

3/19 5/21 7/18

1/14

4/24

7/8

Fig.3 Change with Time of Result for Algae

Strength of porous concrete was close to stable condition
while it was age 28 days.
4) Possibility that coarse aggregate of No.5 and No.7 might
create the preference environment for Ephemeroptera andPtilodactylidae, respectively.
5) Possibility that 18.9% porosity of porous concrete usingcoarse aggregate No.7 would be good adhesion ability against
diatom.
Therefore, porous concrete was appeared to be possibleusing structure that has enough strength, various pores andholes of pores. In addition to, it is considered that thepreference environment for specific creatures is able to becreated for adjustment of aggregate size and porosity.
However, benthos was main target at this experiment. So, thisstudy will be necessary to pay attention to higher order stageof creature for evaluating ecosystem support.

5 ACKNOWLEDMENT
We are grateful to Mr. Mima Masanori, Mr. Masatsugu
Naoki, Mr. Ishikawa Naoya, Mr.Sakai Takehiro, Ms.
Matsuoka Mayu, Mr. Haga Masahiro and members of Water
Use and Environmental Engineering laboratory in Kochi
University, for their kindly cooperation on experiment.
6 REFERENCES
[1]
[2]
[3]
[4]

303

Kawai T, Tanida K, Aquatic insects of Japan : manual with keys andillustrations, Tokaidaigakusyuppankai, 2005..
Maruyama H, Takai M, Tanida K, Illustrated book of aquatic insect
larva. Zenkokunosonkyoikukyokai, 2000.
Yamagishi T, Introduction to the Freshwater Algae, UchidaRokakuho,
1999.
Hori T, An illustrated atlas of the life history of algae, vol. 1,
Uchidarokakuho. 1994.

First International Conference on Geotechnique, Construction Materials
and Environment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7

Alternative Employment of Crushed Shell Particles in Capillary Barrier of Soil
Nakafusa S., Japan Atomic Power Company, Tokyo, Japan
Kobayashi K., Tobishima Corporation, Chiba, Japan
Morii T., Niigata University, Niigata, Japan
Nishimura T., Ashikaga Institute of Technology, Tochigi, JapanABSTRACT: Capillary barrier is a tilting soil layer system which iscomposed of a finer soil layer underlain by a coarser soil layer.
Water which infiltrates into the soil is suspended just above aninterface between the soil layers and flows downward along theinterface, with the result that a vertical movement of water into
deeper soil layers below the interface stops within some length along
the interface. As the water diversion in the capillary barrier isbrought out by a sharp contrast in water retention capacity between
the finer and coarser soils, any material which can provide the sharp
contrast in water retention capacity can be alternatively employed in
the capillary barrier system. A coarse-grained material of crushed
shell particles is selected as an alternative candidate for the lower
layer soil of the capillary barrier, and its water retention capacity ismeasured to examine a practical effectiveness of the crushed shellparticles in the capillary barrier system.

Keywords: Capillary barrier, Water retention capacity,
Crushed shell particles, Fishery byproduct waste
1. INTRODUCTION
Capillary barrier is a tilting soil layer system which iscomposed of a finer soil layer underlain by a coarser soil layer.
Water which infiltrates into the soil is suspended just above
an interface between the soil layers and flows downward
along the interface, with the result that a vertical movement ofwater into deeper soil layers below the interface stops withinsome length along the interface [1]-[5]. Because of thisexcellent diversion of infiltration water, the capillary barrierof soil has been successfully employed in a top cover of waste
landfill. The capillary barrier of soil can be also adopted as
effective measures for slope protection of natural soil or earth
embankment [4]- [6]. As the water diversion in the capillarybarrier is brought out by a sharp contrast in water retention
capacity between the finer and coarser soils, any material
which can provide the sharp contrast in water retention
capacity can be alternatively employed in the capillary barriersystem.
In the study, a coarse-grained material of crushed shellparticles is selected as an alternative candidate for the lower
layer soil of the capillary barrier, and its water retention
capacity is measured to investigate a practical effectivenessof the crushed shell particle. There are two reasons forselecting the crushed shell particles: the coarse-grainedmaterial of crushed shell particles is environmentally friendly
one same as gravel which is usually employed in the capillarybarrier of soil; a better employment of shell will offer somepossible solution to a recycling of fishery byproduct waste.

Firstly geotechnical features of the capillary barrier of soil arebriefly introduced, and some study conducted to determine
the length of water diversion of the capillary barrier of soil isgiven in the paper. Then the crushed shell particles are sieved
into three groups, 9.5-4.75 mm, 4.75-2.0 mm and ones
smaller than 2.0 mm in particle diameter, and their soil-water
characteristic curves (SWCC) are measured by a pressure
membrane method using a SWCC apparatus. An effect ofmass densification of the crushed shell particles caused byeffective confining pressure on the SWCC is also examined.
In constructing the capillary barrier system, an inclusion of
the finer soil particles in the upper layer into the coarser soilin the lower layer should be avoided in order to keep the sharp
contrast in water retention capacity between the finer andcoarser soils. To find a solution to this problem inconstructing the capillary barrier system, lastly, a degree of
the inclusion of sand particles into the crushed shell particles
is investigated by tapping a soil column in the laboratory.
2.

WATER DIVERGENCE BY CAPILLARYBARRIER

2.1 Soil Water Movement in Capillary Barrier System
In the capillary barrier system, water which infiltrates into thesoil is suspended just above the interface between the soil
layers due to a physical difference in the water retention
characteristics of the finer and coarser soils. When theinterface between soils has an inclination as shown
schematically in Fig. 1, the suspended water flows downward
along the interface. This water flow downward along theinterface between soil layers gradually accumulates its massof flow due to continuous infiltration from the soil surface,
and, at some length along the interface, water percolates
Rainfall/ Continuing supply of water

Sand
rfa
alon g inte
Water flowceGravelBreakthroughDiversion length

Fig. 1 Schematic diagram of water flow downward along aninterface between soil layers in capillary barrier system.
305

12.5 20.4

72.0

10.8 21.3+15cm*+10cm+5cm+15cm*+1cm
Veneer board+10cm +10cm * -2cm
+1cm
+5cm
-2cm
+1cm
Ⅰ
-2cm
Sand
°
19.7
t
Ⅱ
e
n
P
P
Ⅲ
vel
face
r
u
s
Gra
Soil

Veneer board

50.0

EC-5

50.0

50.0
Unit of length: cm

46.0

50.0

50.5

Section

Sand soil

Plane* Numeral shows vertical distance (cm) of moisture sensor,
EC-5, embeded below(-)/ above(+) the interface
between soil layers.

Fig. 2 Configuration of capillary barrier soil constructed in
the natural sand slope. Numerals show vertical distances(cm) of the moisture sensor, EC-5, embedded below (-)/
above (+) the interface between sand and gravel.

100

Sand
50

Gravel

0
0.01

0.1

1

10

100

Soil particle diameter, mm

Fig. 3 Grain size distributions of sand and gravel employedin the capillary barrier soil slope.
0.6

Volumetric moisture
content θ, cm3/cm3

2.2 Field Experiments to Determine Water Divergence
The capillary barrier soil was constructed in the natural sandslope as shown in Fig. 2, and volumetric moisture content in
the soil layers was measured together with rainfall intensity
for about four months. Dry densities of the sand layer andgravel layer were 1.38 and 1.67 Mg/m3, respectively. A
completely-permeable polypropylene net was placed over thecompacted gravel layer so that the finer sand particles did not
fall into nor clog void formed in the gravel layer. Inclinationof the interface between the sand and gravel layers was 19.7
degrees in average. Fig. 3 shows grain size distribution curvesof sand and gravel. The sand is classified into “Sand” with
less-5 % fine and coarse fractions; the gravel, commerciallyavailable, is siliceous with mean particle size of 5 to 6 mm.
Relationships of negative pore pressure h with the volumetricmoisture content of the sand and gravel were measured by alaboratory soil column test, and are plotted in Fig. 4. The
SWCC’s in the figure are determined using van Genuchten

Percent finer by mass, %vertically into the coarser soil layer. A horizontal distancefrom the beginning of water flow to this percolation or
breakthrough into the coarser soil layer is called a divergencelength, and is one of important design parameters todetermine a structural configuration of capillary barriersystem and to select a suitable combination of the finer andcoarser soils [1], [2]. In this chapter the divergence length ofcapillary barrier is investigated based on a field measurementof soil moisture in slope soil [6] and a series of laboratory soil
box test.

D: drainage, W: wetting
0.4
Sand(D)
0.2

0.01

Sand(W)
Gravel(W)10100

Negative pressure head h, cm

Fig. 4 Soil-water characteristic curves of sand and gravelmeasured by laboratory soil column tests and estimated byusing van Genuchten equation.
equation [8]. It is found in Fig. 4 that an air entry value of thesand, ha, is about 16 cm, and a water entry value of gravel, hw,
about 1 cm.
The volumetric moisture contents measured in the sand andgravel layers are given in Fig. 5. The upper, central and lower
figures show the measurement at the position Ⅰ, Ⅱ and Ⅲ in
the slope shown in Fig. 2, respectively. The precipitation is
given by an inverse bar with the right vertical axis of the
figure. It’s found that the volumetric moisture content
measured in the sand layer increases rapidly after rainfall andthen decreases slowly, in a successive way from shallow to
deep depth in the soil. But in the gravel layer, as opposed tothe sand layer, there can be seen only very small or negligible
change after rainfall event. It may be understood that this isexactly due to the water diversion by the capillary barrier forlong time duration.
2.3 Estimation of Length of Water Divergence
Comparing the upper figure with the central and lower ones inFig. 5, there can be notified some difference in the soil
moisture changes measured in the gravel layer (denoted by“-2” in Fig. 5 which shows the depth of the moisture sensor
embedded as shown in Fig. 2). That is, although the low
moisture content of soil is maintained in the upper position of
the slope during four-month measurement, the soil moisture
contents measured in the central and lower positions of theslope increase with a small amount immediately after rainfall.
It may be possible to think that this difference in the soil
moisture changes indicates the percolation or breakthrough of
the water flow along the interface into the gravel layers.

306

+5

0.1
0

10

20

Precipitation,
mm･min-1

2

+1cm

0.2

1

+5

+10

+10

+15
-2

+15
-2
30

40

50

60
70
80
Time t, days(a) Measured at the upperl position of slope, A

90

100

110

120

12 October, 2007 to 4 February, 2008
0.5

0Central position of slope, Ⅱ

0.4
+1cm

2

0.3
0.2

+10

0.1
0.0

1

+1cm

+10

-2
0

10

20

30

Precipitation,
mm･min-1

Volumetric moisture content
θ, cm3･cm-3

+1cm

0.3

0.0

Volumetric moisture content
θ, cm3･cm-30Upper position of slope,Ⅰ

0.4

-2
40

50

60
70
80
Time t, days(b) Measured at the central position of slope, B

90

100

110

120

12 October, 2007 to 4 February, 2008
0

0.5
0.4

+1cm

0.3
+5
+10
+15

0.2
0.1
0

10

20

1
2

+5
+10
+15

-2
0.0

Lower position of slope, Ⅲ+1cm

Precipitation,
mm･min-1

Volumetric moisture content
θ, cm3･cm-3

12 October, 2007 to 4 February, 2008
0.5

-2
30

40

50

60
70
80
Time t, days(c) Measured at the lower position of slope, C

90

100

110

120

Fig. 5 Volumetric moisture content in the sand and gravel layers and the precipitation measured during 12 October, 2007 to 4
February, 2008. Note that numerals show the vertical distances of the moisture sensor, EC-5, embedded below (-)/ above (+)
the interface between soil layers in cm. Some dotted lines are interpolated linearly because of lack of data.
Based on this observation and taking account of the locationof measurement points shown in Fig. 2, the divergence lengthof the capillary barrier is estimated to be around 100 cm.
Some equations have been proposed to estimate thedivergence length of capillary barrier by several researchers
such as Ross [1], Kung [9], and Steenhuis, Parlange and Kung[7]. Among these, the equation by Steenhuis, Parlange andKung is effectively adaptable [2], [3]. In the case where an
infiltration rate, q, is much smaller than a saturated hydraulicconductivity of sand (the upper finer soil), Ks, the equation of
the divergence length, L, is given by
L





Ks
tan   -1  ha - hw q(1)

where φ is the slope angle of the interface, ha and hw are the air
entry value of sand and the water entry value of gravel,
respectively, as explained in 2.2. α is an exponential constantdescribing the relationship between the negative pressure
head, h, and an unsaturated hydraulic conductivity, K, of sandnear saturation.
Estimating K from the soil characteristic curve of sand in Fig.
4 by van Genuchten equation and plotting K with h, α nearsaturation can be estimated to be 0.084 cm-1 [6]. Introducing
Ks=8.2×10-3 cm/s which was determined by a laboratory
permeability test, φ=19.7 degrees given in Fig. 2, ha=16 cm
and hw=1 cm determined in Fig. 4 into (1) and assuming that q

is approximately equal to an average value of precipitation,
0.5 mm/min, then about 99 cm is calculated as a value of L.
This L corresponds fairly well to the horizontal length of theslope from the most upper position to the central position Ⅱ
where the soil moisture was measured, which is around 100
cm as explained in the preceding paragraph. Fig. 6 shows thisgood comparison of the diversion length observed in the fieldwith one estimated by (1) by using a double circle.
In order to observe directly the breakthrough of water flow
along the interface and to determine the diversion length, aseries of the laboratory soil box test was carried out. The soilbox with an acrylic front panel, 100 cm in length, 50 cm inheight and 5cm in depth, was placed on a horizontal floor andgravel was compacted into the layer of 10 cm thickness, andthen the sand into 10 cm thickness above the gravel layer inthe soil box. The completely-permeable polypropylene net
was placed over the compacted gravel layer as same as in thefield experiment. After gauze sheet was spread over the soil
surface to protect soil erosion by rainfall droplet, one side of
the soil box was lifted so that the interface between sand andgravel layer had the inclination, then rainfall was supplied
onto the soil surface. Rainfall was simulated by emitting
needles attached to the base plate of water reservoir withconstant head of water. To keep the acrylic front panel clean,
the emitting needles were placed so that the rainfall droplet

307

Diversion length estimated, m

2.0

1.5

mold was loaded with matric suction from 0 to about 200
cmH2O step by step. After this drying process, the matricsuction was inversely decreased to 0 cmH2O to follow a
wetting process. During the drying and wetting processes,
water mass drained from the specimen and absorbed into thespecimen, respectively, was successively measured by using
a burette of the SWCC apparatus. The SWCC’s of the crushed
shell particles A, B and C are given in Fig. 8. Dry densities ofthe specimen of crushed shell particles A, B and C were 1.00,
1.19 and 1.27 Mg/m3, respectively. It’s found in Fig. 8 that

◎ Field observation●Laboratoy soil box tests○ Walter, et al. (2000)

1.0

0.5

0.0
0.0

0.5

1.0

1.5

2.0

Diversion length measured, mFig. 6 Comparison of the diversion length measure in thefield and the laboratory with one estimated by Steenhuis,Parlange and Kung [7].

3.1 Water Retention Characteristics of Crushed Shell
Particles
As the water diversion in the capillary barrier is brought out
by a sharp contrast in water retention capacity between thefiner and coarser soils, any material which can provide the
sharp contrast in water retention capacity may bealternatively employed in the capillary barrier system. In thischapter, a coarse-grained material of the crushed shellparticles is selected as an alternative candidate for the lower
layer soil of the capillary barrier, and its water retention
capacity is measured to examine a practical effectiveness ofthe crushed shell particles in the capillary barrier system.
Being washed and oven-dried, the shells of clams, Meretrix
lusoria and Ruditapes philippinarum, were crushed andsieved into three groups, that is A: 9.5-4.75 mm, B: 4.75-2.0
mm and C: ones smaller than 2.0 mm in particle diameter, as
shown in Fig. 7. The SWCC’s of three groups of the crushed
shell particles were measured by the pressure membrane
technique using a thin porous membrane [10] in the SWCC
apparatus. The crushed shell particles were compacted into asteel mold and saturated with de-aired water. Then aspecimen of the crushed shell particles compacted in the steel
308

Matric suction, cmH2O

ALTERNATIVE EMPLOYMENT OF CRUSHEDSHELL PARTICLESC

Fig. 7 Crushed shell particles sieved into A: 9.5-4.75 mm,
B: 4.75-2.0 mm and C: ones smaller than 2.0 mm inparticle diameter.
50〇: Drying, ●: Wetting

40(a) Particles 9.5-4.75mm

30
20
10
0
0.0

0.1

0.2

0.3

0.4

0.5

0.6

Volumetric moisture contentθ, cm3/cm3
Matric suction, cmH2O

3

B

A

50
40

(b) Particles 4.75-2.0mm

30
20
10
0
0.0

0.1

0.2

0.3

0.4

0.5

0.6

Volumetric moisture contentθ, cm3/cm3
Matric suction, cmH2O

did not fall along the front panel. Intensity of rainfall was
regulated by adjusting the head of water in the water
reservoir.
The divergence lengths which were observed under different
condition of the inclination and the rainfall intensity arecompared with ones estimated by (1) in Fig. 6 by using black
circles. Laboratory results published by Walter, et al. [2] arealso plotted by single circles. It may be found that (1) can bewell employed to estimate the diversion length of capillarybarrier. Because it is known that the diversion length reaches
theoretically as long as 5 to 50 m under a proper combinationof soil particle size [1], a continuing examination in the rangeof larger diversion length will be required.

5040(c) Particles smaller than 2mm

30
20
10
0
0.0

0.1

0.2

0.3

0.4

0.5

0.6

Volumetric moisture contentθ, cm3/cm3

Fig. 8 Water retention characteristics of the crushed shellparticles measured by the pressure membrane method.

Matric suction, cmH2O

50

crushed shell particles as the alternative of gravel usuallyused in the capillary barrier system.

〇: Drying, ●: Wetting

40
30
20
10
0
0.0

0.1

0.2

0.3

0.4

0.5

0.6

Volumetric moisture contentθ, cm3/cm3

Percent finer by mass, %Fig. 9 Water retention characteristics of the crushed shellparticles smaller than 2 mm in particle diameter under theconfining pressure of 50 kPa. Dry density of the specimen
was 1.37 Mg/m3.
100○: before the SWCC test△: after the SWCC test
under confining prssure
50 50kPa.

0
0.01

0.1

1

10

100

Soil particle diameter, mm

Fig. 10 Grain size distributions of the crushed shellparticles smaller than 2 mm in particle diameter measured
before and after the SWCC test under the confiningpressure of 50 kPa.
the water entry value of the crushed shell particles issufficiently smaller than that of gravel in Fig. 4 even in thegroup C, the crushed shell particles smaller than 2 mm inparticle diameter. As shown in (1), the smaller the water entry
value is, the larger the divergence length becomes. This leads
to a practical evaluation that the crushed shell particles can bepreferably adopted as the alternative of gravel usuallyemployed in the capillary barrier system.
Because the coarser material is overlain by the finer material
in the capillary barrier system, a possible increase of dry
density of the coarser layer may cause a change of its water
retention characteristics. Fig. 9 shows the SWCC of thecrushed shell particles C measured under the confiningpressure of 50 kPa which corresponds to about 3 to 4 m
thickness of the upper sand layer. There can be seen, in Fig. 9,
some small increase of the water entry value. As there is not
any change in the grain size distribution of the crushed shellparticle before and after the measurement of the SWCC asshown in Fig. 10, it may be thought that the small increases in
the water entry value of the crushed shell particles under theconfining pressure were caused by mass densification of thespecimens. In Fig. 9, the water entry values of the crushed
shell particles are found to be about 1 to 2 cm. These aresufficiently small comparatively with that of gravel shown in
Fig. 4, which does not reject a practical adoption of the

3.2 Sand Particle Inclusion into Crushed Shell Particles
In constructing the capillary barrier system, an inclusion of
the finer soil particles in the upper layer into the lower coarser
soil layer should be avoided to keep the contrast of water
retention characteristic between both the layer soils sharp.
Although a chemical sheet was employed for a technicalconvenience in the field experiment described in thepreceding chapter, its use should be avoided in constructing
the capillary barrier system from an environmentallyconscious point of view as well as a long-term durability. To
investigate the effectiveness of the crushed shell particles as
the effective alternatives of gravel, the degree of inclusion ofsand particles into the crushed shell particles were examinedby a tapping soil column test in the laboratory. Two
combinations of gravel overlain by the sand and the crushed
shell particles overlain by the sand were tested and compared.
The grain size distributions of the sand and gravel are given in
Fig. 3. The group C of the crushed shell particles was selectedin the test because of the preferable reason mentioned in thepreceding section, and its grain size distribution is shown in
Fig. 10.
Firstly gravel and the crushed shell particles were poured andlightly compacted into two 1,000cc-graduated glass cylinders
35cm in height and 6.8cm in inner diameter to form the sameheight of cylindrical volume as shown in Fig. 11(a). Then thesand was poured above the cylindrical volumes of gravel andthe crushed shell particles so both the heights of thecylindrical volume in the graduated glass cylinder reach thesame. After being placed on a horizontal desk, the graduatedglass cylinder was tapped around its base cyclically,
continuously and lightly by a wood hammer with horizontal
amplitude of 5 cm. Fig. 11(b) shows the cylindrical volumes
in the graduated glass cylinders after the cyclic tap 100 times.
A higher line given in Fig. 11(b) shows the initial height ofthe cylindrical volume before the cyclic tap. In the case ofgravel overlain by the sand, the left-hand side in Fig. 11(b), a
large amount of the sand particles fell down into the lower
volume of gravel, and the initial height of the cylindricalvolume lowered to a position denoted by a dotted line.
Contrary to this, the cylindrical volume of the crushed shellparticles overlain by the sand keep its initial height almost thesame after the cyclic tap 100 times, and the interface between
the sand and the crushed shell particles was clearly observed.
Both of these observations suggest less inclusion in the caseof the crushed shell particles overlain by the sand than in thecase of gravel overlain by the sand. Although the grain size ofthe crushed shell particles C is smaller than gravel as shownin Fig. 3 and Fig. 10, the water entry value of the crushed shellparticles is very small compared to gravel as shown in Fig. 8
and Fig. 9 and, furthermore, small inclusion of the finer soil
particles from the upper layer in constructing the capillarybarrier system can be well expected as observed above. It may
be thought that these special features make the crushed shellparticles attractive as the alternative of gravel in the capillary309

by Steenhuis, Parlange and Kung [7].
Because the coarse-grained material of the crushed shellparticles is the environmentally-conscious one same as gravel
which is usually employed in the capillary barrier system anda better employment of shells will offer some possible
solution to the recycling of fishery byproduct waste, theeffectiveness of the crushed shell particles as the alternativeof gravel was investigated by a series of the pressure
membrane test and the tapping soil column test in thelaboratory. It was found that the water entry value of thecrushed shell particles smaller than 2.0 mm in particle
diameter was extremely small even under the confiningpressure. It’s also found in a series of the tapping soil column
test that the inclusion of the sand particles in the upper layer
into the lower layer could be well eliminated by using thecrushed shell particles. As the inclusion of the sand particles
into the lower coarse-grained layer is one of important
problems to be solved in constructing the capillary barriersystem, the result of the tapping soil column test will support
attractive employment of the crushed shell particles in thecapillary barrier system.

(a) Gravel (left) and the crushed shell particles (right)
poured and lightly compacted in the glass cylinder. A
solid line shows the height of both the lower cylindricalvolumes.

5 ACKNOWLEDGMENTS
The present study is supported by the Grant-in-Aids for
Scientific Research (B), No. 22380127, and Scientific
Research (C), No. 22560507, made by the Ministry of
Education, Science, Sports and Culture of Japan.
6[1](b) Cylindrical volumes in the glass cylinders after thecyclic tap 100 times. The left-hand side is gravel overlain
by sand, and the right-hand side the crushed shell particles
overlain by sand. A solid line shows an initial height, anddotted lines show the height after the cyclic tap.
Fig. 11 Inclusion of the sand particles in the upper layer
into gravel and the crushed shell particles in the lower
layer observed in the tapping soil column test.
barrier system.
4

CONCLUSIONS

Geotechnical features of the capillary barrier of soil werebriefly introduced, and the study conducted to determine thelength of water diversion of the capillary barrier was
summarized in the paper. Soil moisture changes in the gravel
layer were well compared those in the sand layer to confirm apractically excellent divergence of infiltration water along thetilting interface between the sand and gravel layers. The
divergence length of the water flow along the interface wasestimated using the soil properties determined by thelaboratory tests and the structural configuration of thecapillary barrier of soil. Fairly good correspondence of thedivergence length between the observation and the estimation
indicated a practical effectiveness of the equation proposed

REFERENCES

Ross B., “The diversion capacity of capillary barriers,” Water
Resources Research, 26(10), 1990, pp. 2625-2629.
[2] Walter M. T., Kim J.-S., Steenhuis T. S., Parlange J.-Y., Heilig A.,
Braddock R. D., Selker J. S. and Boll J., “Funneled flow mechanisms ina sloping layered soil: Laboratory investigation,” Water Resources
Research, 36(4), 2000, pp. 841-849.
[3] Smesrud J. K. and Selker J. S., “Effect of soil-particle size contrast on
capillary barrier performance,” Journal of Geotechnical andGeoenvironmental Engineering, 127(10), 2001, pp. 885-888.
[4] Tami D., Rahardjo H., Leong E.-C. and Fredlund D. G., “A physicalmodel for sloping capillary barriers,” Geotechnical Testing Journal,
27(2), 2004, pp. 173-183.
[5] Rahardjo H., Krisdani H. and Leong E.-C. “Application of unsaturated
soil mechanics in capillary barrier system,” Proceedings of the Third
Asian Conference on Unsaturated Soils, Nanjing, China, 2007, pp.
127-137.
[6] Morii T., Takeshita Y., Inoue M. and Matsumoto S., “Alternative
measures for soil slope stability using capillary barrier of soil,”
Proceedings of the Fourth Asia-Pacific Conference on Unsaturated
Soils, Newcastle, Australia, 2009, pp. 319-324.
[7] Steenhuis T. S., Parlange J.-Y. and Kung K.-J., “Comment on ‘The
diversion capacity of capillary barriers’ by Benjamin Ross,” Water
Resources Research, 27(8), 1991, pp. 2155-2156.
[8] Stephens D. B., “Characterizing hydraulic properties,” Vadose Zone
Hydrology, CRC Press, Inc., 1996, pp. 183-187.
[9] Kung K.-J. S., “Preferential flow in a sandy vadose soil 2, Mechanism
and implications,” Geoderma, 46, 1990, pp. 59-71.
[10] Nishimura T., Toyota T. and Koseki J., “Evaluation of apparent
cohesion of an unsaturated soil,” Proceedings of the FourthAsia-Pacific Conference on Unsaturated Soils, Newcastle, Australia,
2009, pp. 109-114.

310

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

A Novel Statistical Approach for Investigating the Significant Factors that
Influence the Performance Score of the Turkish Method
Kshama S. Roy, Faculty of Engineering and Applied Science, Memorial University of Newfoundland, Canada
Kamrul Islam, Civil and Environmental Engineering, University of Alberta, Canada
Md. Shams Arafat, Department of Civil Engineering, Bangladesh University of Engineering and Technology,
Dhaka, Bangladesh.

structural system, structural and non-structuralcomponents, and structural performance [2-4].
Vulnerability assessment using those proceduresshow accurate and realistic performance of a building
if the building was built according to proper building
design code and architectural features. Recent years,
Bangladesh has seen the construction boom of multistoried (mainly 6 stories and above) reinforced
concrete frame (RCF) buildings in the capital city of
Dhaka where standard building codes are notfollowed properly. Most of these buildings have open
parking space in the ground floors resulting in softstory construction. Some of them have structural
irregularities. Seismic vulnerability assessment ofexisting buildings of Dhaka city is, therefore, ofmajor importance. According to the opinion of a
German earthquake expert (Dorka, 2005) of the
University of Kassel, Germany, Turkish method ofseismic risk assessment may be applicable for adeveloping country like Bangladesh [5]. The Turkishmethod, based on the statistical analysis of observed
earthquake damages of RCF buildings in Turkey andrelated building attributes is expected to provide
reasonable results for assessment of buildings in
Bangladesh [6, 7]. Turkish method is a two level
seismic risk assessment procedure considering asignificant number of factors [6]. It gives aperformance score for each building being assessedconsidering the factors. But as it uses a lot of
empirical equations and tables for determining thescore, it is difficult to find the most significant factorsthat affect the score. The main aim of this study is todevelop a statistical approach using Design ofExperiment (DOE) methodology to investigate thesignificant factors of Turkish method described by
Ozcebe,G. et al. (2003) [6] and to develop a linear
regression equation for determining the performance
score to compare with the results given by theTurkish Method.

ABSTRACT: This article addresses a statistical approachto evaluate the effectiveness of Turkish Method for seismic
vulnerability assessment of low to medium rise reinforced
concrete frame (RCF) buildings. Design of Experiment(DOE) methodology is extensively used to investigate themost important factors that affect the performance score
given by the Turkish method using non-commercial
software Design Expert 7.1.3. A 210-3 Fraction Factorial
Design is used to determine the factors that significantly
influence the performance score. The results show thatpounding effect, topographic effect, plan irregularity, areaof the ground floor, number of frames in X and Y direction
and the interaction between the number of frames in X andY direction are the most significant factors. Finally a linear
regression model is developed for computing theperformance score to compare with the performance score
given by the Turkish Method.

Keywords: Fractional Factorial Design, DOE,
Turkish Method, Seismic Vulnerability.
1. INTRODUCTION
Recent years the world has seen some severe threats
imposed by several earthquakes in different countries.
Those earthquake events have caused serious
disruption in normal way of livings including deaths,
economic loses and damage of infrastructures. Most
recent earthquakes in Japan (2011), New Zealand(2010, 2011), Mexico (2011), Indonesia (2010, 2011),
China (2010), and Haiti (2010) have brought
attention to researchers to predict the severity ofdamage of infrastructures [1]. Many cities of thedifferent countries of the world contain a significant
number of vulnerable buildings under considerable
seismic risk. Seismic risk assessments of thosebuildings are gaining more attention to assesscorrectly the degree of risk to which such criticalstructures may be subjected to in the event of anearthquake. Various vulnerability assessment
procedures are found in the literature while those
procedures are based on building plan, foundation,

311

DOE techniques for experimentation which is used inthis study.

2. TURKISH METHODThe Turkish seismic vulnerability assessment method
for buildings consists of two-level evaluation: (i)level one – walk down evaluation and (ii) level two measurements at ground level and basement.

4. EXPERIMENTAL PROCEDURE
4.1 Strategy of experimentation
For performance score of the building, basically
Level 2 survey is important as it also incorporates theLevel 1 survey result. So to find the factors that affectthe performance score, factors for Level 2 survey arechosen as factors for DOE and the response, theperformance scores are found by giving the factors as
inputs in the program. The levels of the factors arechosen on the basis of the survey that was done in theward 75, 76 and 77 of the old town of Dhaka citywhich is the capital of Bangladesh [8]. The runs forthe fractional factorial design are run by theMicrosoft Excel software program and then the data
are analysed by the non-commercial software, Design
Expert 7.1.3.

The walk down evaluation procedure consists oftaking notes of different external features of thebuilding. Buildings are then described as belonging
to high risk, moderate risk or low risk. After level 1
survey, buildings falling into the moderate and highrisk levels can be subjected to more detailed level 2
surveys to determine their performance scores.
Details of Level 1 and Level 2 Survey procedures canbe found in [8]-[9]. Once the vulnerability parametersof a building are obtained from two-level surveys andits location is determined, the seismic performance
scores for survey levels 1 and 2 are then calculated(Sucuoglu et al., 2003) [10].
A general equation for calculating the seismic
performance score (PS) can be formulated as follows:

4.2 Factors and Response of Interest and theirlevelsPS = (Initial Score) - Σ (Vulnerability parameter) x(Vulnerability Score)Ten (10) parameters of the Level 2 survey of Turkishmethod are used as factors here. The factors are A)
Pounding effect, B) Topographic effect, C) Planirregularity, D) Area of the ground floor, E) Area of
the above floor, F) No of frames in X direction, G)
No of frames in Y direction, H) Column dimension_x, I) Column dimension _y and J) No of columns.
Performance score is used as Response. Among theten factors A) Pounding effect, B) Topographic effectand C) Plan irregularity are considered as categoricalfactors and others are considered as numeric factors.
For the categorical factors the high and low levels are“yes” and “no” respectively. But for the other factors
the levels are shown in Table 1.

The detailed procedures for vulnerability scoring are
given by Sucuoglu et al., 2003 [10].
3.

DOE METHODOLOGYDesign of experiment (DOE) and statisticaltechniques are widely used to optimize processparameters and to develop a mathematical
relationship between the input parameters and theoutput variables. It involves the study of any given
system by a set of independent variables (factors)
over a specific region of interest (levels) and provides
a straightforward technique with linear graphs todetermine the relationship (interactions) between theconsidered factors, which can be used for practicalexperimentation. By using DOE, it is possible toinvestigate the experimental process, to screen theimportant variables (or factors), to build amathematical model with prediction and even to
optimize the responses where necessary. Among thedifferent techniques in design of experiment method,
Fractional factorial design is one of the common

4.3 Experimental Design
As ten (10) factors and two levels are considered, 210= 1024 runs were required if full factorial design wasused. So to minimize the runs, 210-3 fractional
factorial design is used. So 210-3 =128 runs are
required for the experiment for finding the importantfactors that affect the performance score. Thoughseven effects are sacrificed for doing the 210-3
fractional factorial design, the main factor and the312two factor interactions are kept free from aliases. So
it is a Resolution V design that means two factors

5.1 Statistical Testing (ANOVA analysis)Table I: Levels for Numeric Factors
Low ActualHigh Actual

FactorName

D

Area of the ground floor

500 ft2

Area of the above floor

2E

400 ftHigh Coded

3000 ft2

Low
Coded
-1.0

5000 ft2

-1.0

+1.0
+1.0

+1.0

F

No of frames in X direction

2

20

-1.0

G

No of frames in Y direction

2

20

-1.0

+1.0

H

Column dimension _x

3 in

40 in

-1.0

+1.0

I

Column dimension _y

3 in

40 in

-1.0

+1.0

J

No of columns

4

50

-1.0

+1.0

The name ANOVA (Analysis of variance) stems
from a partitioning of the total variability in the
response variable into components that are consistent
with a model for the experiment. ANOVA is used todetermine the significant factors and their interaction.
The model was checked (checking the assumption of
ANOVA, the R2 and adjusted R2 and also lack of fit)
and found that model with natural log transformation
gives the higher R2 as well as high insignificant lackof fit values. So the results given here are the results
obtained after natural log transformation. The
ANOVA table is given in the Fig. 1.

interactions are aliased with three factors interaction
and single factor is aliased with four factors
interaction. However “Sparsity of effects” principle isused, so three factors interactions and above areconsidered negligible
5. ANALYSIS OF RESULTS
After giving the input data in the Design-Expert 7.1.3,
the significant effects and interactions are determinedby using Effects list and half-normal plot. From theeffects list, factors A, B, C, D, F, G and theinteraction between F and G are found having
maximum contribution. So these are used as models.

The calculation in Fig 1shows that MS (treatment) islarger than MS (residuals) which results a larger F

Fig. 1: ANOVA table for selected factors

313

In the Fig. 4, most of the data are randomly scatteredwhich indicates the sequence of run was random. In
Fig. 5, the plot shows that the data points are split
evenly by the 450 line means no transformation. It
also shows that the model provides accurate values
even though most of the Effects are discarded.

value (>> 1) (test static). Moreover p-value is <<0.05 for all the factors (A, B, C, D, F, G) and also for
the interaction of F & G. So all the selected effects
are highly significant which indicates that the six
factors, A, B, C, D, F, G and the interaction FG havethe significant effect on the response (performance
score).
Again the value of R-squared is 0.9188 which is near
to 1. That means the model is good enough and91.88% of variability of data can be explained by themodel. The "Pred R-Squared" of 0.9072 is inreasonable agreement with the "Adj R-Squared" of
0.9139. "Adeq Precision" measures the signal tonoise ratio. A ratio greater than 4 is desirable. In thismodel, the ratio of 52.918 indicates an adequate
signal. So, this model can be used to navigate thedesign space.

5.3.3 Box-Cox plot
In the Box-Cox plot, the current line (blue line) is
between the ranges (the red lines) and near the bestvalue line (the green line). This recommends for no
transformation. So, we can conclude that theassumptions of ANOVA are satisfied.
5.4 Regression Model (Model Equation)
The model equation is given in terms of coded unit in
the Fig7. In the equation 4.12 is the overall averageof the effects and 0.026, 0.022, 0.042, 0.038, 0.084,
0.084 and 0.045 are the effects of A, B, C, D, F, G
and FG consecutively in terms of coded factors. So
from the above equation in coded factors, it can beseen that F and G have the largest effect on
performance score.

5.2 Assumption of ANOVAa) All samples are random samples from theirrespective populations.
b) All samples are independent of one another.
c) Departures from group mean are normally
distributed for all groups.
d) All groups have equal variance.

5.5 Model Graphs and Interpretation

5.3 Checking the Adequacy of the Model
5.3.1 Normality and Constant Variance
Assumption

The Fig. 8 shows the effect of F on performance
score when all others are fixed at high level. Thisshows that F has significant effect on score and thescore increases gradually as the F increases from 2 to
20. The Fig. 9 shows the effect of G on score when
all others are fixed at high level. This shows that G
has a significant effect on score. As the G increases
from 2 to 20, the score also increases significantly. In
both Fig. 10 and Fig. 11, interaction between F and G
are showed when all other factors are at low level(Fig. 10) and when all others are at high level (Fig.
11). In the Fig. 14, the score (response) increases
slightly with the changes of F from 2 to 20 when G is
low (2), the red line. But score increases significantlywith the changes of F from 2 to 20 when G is high(20), the black line. Again in the Fig. 11, scoreincreases very slightly with the changes of F from 2
to 20 when G is low (2), the red line. But scoreincreases significantly with the changes of F from 2

To verify the normality assumption, that the error is
normally distributed, we can use a normal probability
plot of the standardized residuals as shown in the Fig.
2 and Fig. 3.
The normal plot of residuals shows that except somepoints, most of points follow a straight line whichmeans that the residuals are normally distributed. The
points are following an S –shape curve after thetransformation. So we have to check the otherresidual plots. From Fig. 2, the plot of Residuals Vs
Predicted looks like well scattered. The randomly
scattered plot indicates the constant range of residuals
across the graph.
5.3.2 Randomness and Checking for Goodness-ofFit314

to 20 when G is high (20), the black line. Also it isclear that the performance score (response) is also
higher when all the other factors are at low level.
The cube plot in Fig 12 also shows that performance
score is maximum when A, B, C, D are minimum and
F and G are maximum.

315

Fig. 2: Normal Plot of Residuals

Fig. 3: Residuals vs Predicted

Fig. 4: Residuals vs. Run

Fig. 5: Predicted vs. Actual

Fig. 6

Fig. 7316

Fig. 8: Box-Cox Plot

Fig.9: Model equation

compare the performance score of the Turkish
Method to the score obtained from the program.

6.0 MODEL VALIDATION
After developing an appropriate model, several runs
were conducted that were not in the initial runs andeach of the time the performance score found by the
optimization chart was almost similar to theperformance score obtained from the program. This
proves that the model developed can give almost
same result as the program.

The paper also shows the adequacy of the model
equation by various diagnostic plots which are all inagreement with the model equation. The R2 value
obtained from the ANOVA result for this model isalso good which indicates good fitness of the model.
The predicted R2 and adj R2 are also found inreasonable agreement. Finally, few test runs are also
performed to check the accuracy of the optimum
solution and achieved responses are also found to be
in good agreement with predicted response.
8. REFERENCES[1]http://earthquake.usgs.gov/earthquakes/eqinthenews/[2]Hassan, A.F., Sozen, M.A. Seimic vulnerality
assessment of low-rise buildings in region of infrequent
earthquake. ACI Structural journal 1997; 94(1): 31-39[3] Gulkan, P.and Sozen, M.A. Procedure for determining
seismic vulnerability of building structures. ACI Structural
journal 1997; 96(3): 336-342[4] Kircher,C. Reitherman, R.K., Whitman, R.V., andArnold,C. Estimation of Earthquake lossess to buildings.
Earthquake Spectra, EERI 1997; 13(4): 703-720[5] Dorka, U. (2005) Personal Communication.
[6] Ozcebe, G., Yucemen, M. S., Aydogan, V., and A.
Yakut (2003) “Preliminary Seismic Vulnerability
Assessment of Existing Reinforced Concrete Buildings in
Turkey- Part I: Statistical Model Based on Structural
Characteristics, Seismic Assessment and Rehabilitation of
Existing Buildings, NATO Science Series IV/29, 29-42.
[7] Ahmed, M. Z., Islam, M. K., Ahsan, R. and Ozcebe, G(2007) " Seismic Vulnerabilty Assessment Methods: A
Possible Application to Bangladesh", First Bangladesh
Regional Science Association (BRSA) Conference, Dhaka.
[8] Ahmed, M.Z., Islam, K., Roy, K. S., Arafat, M. S. and
Al-Hussaini, T.M. (2010) “Seismic Vulnerability
Assessment of RCF Buildings in Dhaka City”, in Proc. 3rd
Int. Earthquake Symposium, Dhaka, Bangladesh.
[9] Roy, K. S., Islam, K. and Arafat, M. S. (2010) “Seismic
Risk Assessment of Existing Low-rise Buildings in
Unplanned Urban Regions of Dhaka City” in Proc. 2nd Int.
Conf. on Construction in Developing Countries (ICCIDC2), Cairo, Egypt.
[10] Sucuoglu, H. and Yazgan, U. (2003) "Simple Survey
Procedures for Seismic Risk Assessment In Urban Building
Stocks", Seismic Assessment and Rehabilitation of Existing
Buildings, 97-118, NATO Science Series, IV/29, Editors:
S.T. Wasti and G. Ozcebe, Kluwer.

Fig. 10: Interaction graph between F and G

Fig. 11: Interaction graph between F and G

7.0. CONCLUSION
The paper gives a linear model to calculate theperformance score of the Turkish method. It also
describes clearly that Pounding effect, Topographic
effect, Plan irregularity, Area of the ground floor,
Number of frames in X and Y direction and theinteraction between the number of frames in X and Y
directions are the most significant factors for
performance score of the Turkish Method. The model
equation developed by ANOVA can be used to317

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Ultimate Strength and Fatigue Life of Concrete Structural Elements
Reinforced with Carbon Fiber SheetHIDENORI Tanaka, Environmental and Urban Engineering, Gunma National College of Technology
YUTAKA Toi, Institute of Industrial Science, University of TokyoABSTRACT:Continuum Damage Mechanics(CDM) was applied toevaluate the mechanical behavior of concrete structural elements
reinforced with carbon fiber sheet(CFsheet) under quasi-static andcyclic loading. A constitutive equation for elasto-plastic damageable
solids is formulated by using Drucker-Prager’s equivalent stress andidentified by the uniaxial compressive and tensile experimental
results for concrete , tensile results for CFsheet. And then finite
element analyses are carried out for real-scale cantilever reinforced
concrete slabs and concrete blocks with CFsheet. Finally It was
concluded that the application of CDM for concrete structuralultimate strength and fatigue durability comparing the experimental
data and the calculation by 2-dimensional finite element method( FEM) was very useful.

Keywords: Ultimate strength, Fiber sheet, Fatigue life
Concrete, Damage Mechanics
1. INTRODUCTION
Concrete is a standard brittle material in civil engineering.
Reinforced concrete structures with steel bar resisting tension
have been used for the infrastructures such as buildings,
bridges and tunnels, due to its durability and economy. In
recent years, special attention has been paid to CF sheets toresist against earthquake repair damage and prevent
separation failure[1]-[3].The CF sheets is of light weight(1/5of steel),high strength (10 times of steel),high rigidity andhigh durability as well as easy to construct. Considering
future development of CF sheets, it is important to establish
the analytical evaluation of the strength and durability ofconcrete structural members reinforced with CF sheets.
The elasto-plastic model, the smeared crack model and thediscrete crack model considering cracking in concrete havebeen proposed as the analytical model for reinforced concrete(RC) structures[4]. On the other hand, there has been muchprogress in the application of CDM[5] to metal materials. The
mechanical deterioration is represented by the internal statevariable D called damage variable and the strain energy
release rate Y conjugate with D in damage mechanics. It is
expected as a branch of mechanics essentially applicable tofatigue fracture and residual life prediction as well as staticand dynamic strength analysis at the material test andstructure level. The validity of the application of damage
mechanics to concrete structures has been recognized throughthe researches on reinforced concrete rigid frames[6],shortfiber-reinforced concrete[7] and high-strength concrete[8].
The development of the analysis program based on damage
mechanics and its experimental validations are conducted in

the present study for the purpose of establishing the method
for ultimate strength and lifetime evaluation of the concrete
structuralmembers reinforced with CFsheet. In theapplication of damage mechanics models to concrete and the
CFsheet, the two-dimensional elasto-plastic damage
constitutive equation is formulated by using Drucker-Prager’s
equivalent stress in order to consider the difference between
the tensile and the compressive strength of concrete. The
formulated constitutive equation is implemented in thetwo-dimensional finite element program. The material
constants are identified by using the material test results
under uniaxial tension and compression and applied to theultimate strength and fatigue life analysis.
2. FORMULATION AND IDENTIFICATION OF
ELASTO-PLASTIC DAMAGE CONSTITUTIVEEQUATION
2.1 Elasto-Plastic Damage Constitutive Equation
The dissipation potential for the growth of plastic strains,which is the sum of the plastic potential and the damage
potential is expressed by the following equation:

F  F p  ,  ; D  FD Y ; p , D 

Y 
S1
 
  eq     y 
S 2  11  D   S1 

S 2 1(1)

Where F p is the potential for the growth of plasticstrains,which is a function of the effective stress  ,theplastic hardening parameter  and the scalar damage
variable D . FD is the potential for the evolution of damage,which is a function of the strain energy release rate Y ,theequivalent plastic strain p and the damage variable D .
In the formulation of the constitutive equation, the yieldfunction is assumed as follows:f  F p   eq     y  0 eq   eq 1  D 

 eq   I1  J 2' 

(2)
(3)

1/ 2

(4)
Where the following notations are used:  eq ;effective
Drucker-Prager’s equivalent stress,  y ;the yieldstress,  ;the material parameter, I 1 ;the first invariant ofstress and J 2' ;the second invariant of deviatoric stresses.
The yield function given in equation(2) is assumed as theplastic potential.

317

Specimen

Finally, The relation between the effective stress increment
and the strain increment were derived [9],[10]:d   C d

Displacement
measure equipment(5)

Where C is the effective stress-strain matrix considering
the elasto-plastic damage. The relation between the stressincrement and the strain increment is expressed by thefollowing equation:(6)d  1  D d    dD  D d

Fig.1 Displacement measure equipment (for compression),,

EXPERIMENT

MODEL

-6000

-4000

-2000

0.0
-10000epd-8000

Where Depd is the tangential stress-strain matrix considering
the elasto-plastic damage, which relates the stress incrementwith the strain increment.The damage evolution equation as
given by the following equation is used in the present
study[5].

0
-10.0

-20.0

S2

Y 
dD    dp
 S1 

-30.0(7)Where S1 and S 2 are the material constants. It is assumed
that the damage evolves with an increase of the equivalent
plastic strain. The strain energy release rate Y is a function ofYoung’s modulus E and equivalent stress  eq as
expressed by the following equation:

Fig.2 Compressive stress-strain curves for concrete,

2,

EXPERIMENT

MODEL

4.0

2 E 1  D 

2

(8)
3.0

2.2 Identification of Material Patrameters
The material constants of concrete used in the constitutive
equation have been determined as shown in Table 1 by the
curve-fitting technique for the uniaxial compressive test
results given by displacement measuring method (Fig.1) andtensile results by using strain gauge method. In this study, theplus sign means tensile and minus means compressive
respectively. Fig.2 and Fig.3 show the stress-strain curves ofcompression and tension. In Fig.2, one of the stress-strain
curves for experiment sharply dropped after peak. It would
be occurred by shear fracture due to eccentric loading after
peak.  pd and Dcr in Table 1 are the damage plastic strain
threshold and the critical damage at crack initiation
respectively. When the equivalent strain exceeds  pd , thedamage evolves according to equation (7). And the damage
variable reaches Dcr ,the material fractures and the stress isreleased. The experimental results were shown until thetensile strength in Fig.3. The post peak behavior after tensile
strength could not be obtained because of brittle fracture ofconcrete and strain gauge’s breaking.
The material constants of the CFsheet has been determined
as shown in Table 2 from the tensile test result given in Fig.4.
318

STRESS (MPa)

 eq

-40.0

STRAIN （μ）

Break

Breakag

2.0

1.0

0.0

0

50

100

150

200

250

300

STRAIN (μ)

Fig.3 Tensile stress-strain curves for concrete
MODEL

EXPERIMENT

2,000

1,500

Break
Breakage
STRESS （MPa）

Y

STRESS (MPa)TFp  Fp 
 C



    d C 1 
T


F
F
 H   p  C p  eq FD   

 1  D Y 



1,000

500

0
0

1,000

2,000

3,000
STRAIN (μ)

4,000

Fig.4 Tensile stress-strain curves for CFsheet

5,000Table 1 Material constants for concrete
E (Young’s modulus)

29.7×103 MPa

ν (Poisson’s ratio)

0.17

α (equivalent stress)

0.72

σy (yield stress)

0.75

K (plastic hardening)

40.0 MPan (plastic hardening)

0.215

Sp1 (damage parameter)

0.215×10-3 MPa

The 2-dimensional FEM model for this structure is describedin Fig.6,and the damage(cracking ) distribution by numericalanalysis is shown in Fig.7.In this Fig., The cracking was
initiated at the static loading P=32.8(kN).

MPa

Sp2 (damage parameter)

1.55

εpd (damage threshold)0.00

Dcr (critical damage)

2.75×10-5

Fig.6 FEM model for clamped RC structure (unit;mm)

Table 2 Material constants for CFsheet

P=32.8(kN)

P=47.4(kN)

3. STATIC AND CYCLIC LOADING TESTS FOR RC
STRUCTURAL MEMBERS

P=82.5(kN)

The real size clamped RC Slab structure used in theexperiments is schematically shown in Fig.5.
Vertical loading

P=31.6(kN)

Carbon Fiber Sheet

Fig.7 Cracking distribution by FEM
Steel bar

Fig.5 Test specimen(unit;mm)
This structure is a part of the bridge which has been
constructed in the early of 1970’s. It needs some repair andreinforce to use as a bridge. The static and cyclic loading test
as a traffic load was carried out, and the ultimate strength andthe affect of the strength after 2-millions cyclic loading were
evaluated from this test and numerical analysis.
319

Fig.8 Specimen before test

4. ADHESIVE FRACTURE OF CONCRETE BLOCK
REINFORCED WITH CFSHEETIn this Section, The adhesive fracture mode such as
de-lamination , surface detachment between the CFsheet,epoxy resin and concrete were discussed.
4.1 A New Constitutive Equation for shear behavior
Most of ultimate strength of concrete structures reinforced
with the CFsheet have been determined by adhesive fracturemode between the CFsheet, epoxy resin and concrete surface.
Therefore, A new equivalent stress was introduced toevaluate the mode as follows[12];CFsheet breaking line

 eq   I1  J 2'     max    max
1/ 2

Fig.9 CFsheet fracture after test
EXPERIMENT
2D-F.E.M

EXPERIMENT (after 2 millions)
2D-F.E.M (after 2 millions)

Where the following notations are used:  and  ;thematerial parameter,;Macaulay’s brackets,  max ;maximum principle stress,  max ; maximum shear stress.
another parameters are same in equation(4).
 and  were determined by the adhesive strength test
shown in Fig.11. In this test, the adhesive length was changed
20,40,80(mm) (refer to Table 3).

100

Sheet Break80

LOAD (kN)

(9)

60

40

20

0
0

5

10

15

20

25

DISPLACEMENT (mm)Fig.10 Relationship between load and displacement320

Fig.11 Adhesive strength test
Table 3 Test cases

CASE

Adhesive
Adhesivelength L (mm) width W (mm)
CASE1
80
80
CASE2
40
80
CASE3
20
80
CASE1 (w=80,L=80)

MAXIMUM LOAD (kN)

Fig.8 and Fig.9 are specimen before and after quasi-static
loading test, respectively. The CFsheet breaking mode wasfound as a final fracture mode. The relationship between loadand displacement at loading- point is shown in Fig.10.The
concrete cracking under the CFsheet as well as the verticalcrack at the root
have corresponded well with theexperimental result as shown in Fig.7 and Fig.9.The numberof vertical cracks at the root is one in the experiment and is
two in the analysis. This difference is probably due to themodeling of the steel bars, which is considered as theequivalent layer area model.
The experimental observation for the load-displacement
curve, in which the brittle fracture behavior with a sudden
load drop takes place after the fracture of the CHsheet, hasalso well been simulated as shown in Fig.10.The calculated
ultimate load of 82.5(kN) has agreed well with theexperimental value of 77.0(kN)[11].
The result of same ultimate load after 2-millions cyclicloading expressed the sine curve which has 28.3(kN) of
amplitude and 2(Hz) of frequency as a design traffic load isalso shown in Fig.10.The calculated ultimate load of
72.7(kN) has agreed well with the experimental value of
61.8(kN). It was found that the ultimate strength due toaffection of pre-cyclic load., which may reduce about 12％
the ultimate strength more than that of intact specimen bynumerical analysis. And the experimental result wasapproximately 20％bigger than the numerical results.

CASE2 (w=80,L=40)

CASE3 (w=80,L=20)

20
15
10
5
0
0

20

40

60

80

ADHESIVE LENGTH (mm)Fig.12 Maximum load and adhesive length100

Table 4 Material property
E (Young’s modulus)

30,000 (MPa)ν(Poisson’s ratio)

0.17α(plastic potential)

0.85

β(plastic potential)

0.035

δ(plastic potential)

0.01

load

Carbon fiber sheet (t=0.14)

σy (yield stress)

1.50 (MPa)σf (fatigue limit stress)

1.875 (MPa)K (plastic hardening)

45.0 (MPa)n (plastic hardening)

0.175

Spf1 (damage parameter)

9.25×10-4 (MPa)

Epoxy resin (t=0.15)

Spf2 (damage parameter)

1.55

Se1 (damage parameter)

4.25×10-3 (MPa)Se2 (damage parameter)

5.375

εpD (damage threshold)

0.00

80

7.45×10-2

Dcr (critical damage)The relationship between maximum load and adhesive
length is shown in Fig.12. The load becomes bigger
corresponding to the adhesive length, but the relation is
nonlinear. It was found that there is a certain effective lengthof the CFsheet to resist the shear deformation.

 Y 

dp , dD  
 S e1 

Se 2

de

(10)

Where S pf 1 and S pf 2 are material constants regarding plastic
fatigue and dp is the increment of equivalent plastic strain.
S e1 and S e 2 are material constants regarding elastic fatigue .
de is the increment of equivalent elastic strain[12].
The CFsheet and epoxy resin were hypothesized as linear
elastic material in this section[12]. The applied cyclicloading ,which has 5(Hz) of frequency for CASE3 and the2-dimensional model for FEM are shown in Fig.13 and Fig.14.
Finally, the material property was shown in Table 4

EXPERIMENTS

5
0
2

4
6
CYCLES

8

10

2

3

4

5

Log N
EXPERIMENTS

2D_FEM

12.0
10.0
8.0
6.0
4.0
2.0
0.0
0

0

1

Fig.15 Relation fatigue fracture number and Pmax(CASE2)

LOAD (kN)

LOAD (kN)

Pmax=8.836

2D_FEM

12.0
10.0
8.0
6.0
4.0
2.0
0.0
0

15
10

100

Fig.14 2-dimensional model for FEM(unit;mm)

LOAD (kN)s pf 2

100

B part4.2 Adhesive Fracture Analysis Under Cyclic Loading
The damage evolution would be hypothesized following
equation. The plastic damage is adopted the former relationof equation(10) and the elastic damage is adopted the later
relation of equation(10),respectively.

 Y dD  
S 
 pf 1 Concrete prism

A part

1

2

3

4

Log N
Fig.16 Relation fatigue fracture number and Pmax(CASE3)

Fig.13 The applied cyclic loading (for CASE3)
321

5

quantitative evaluation of the chemical aging, themechanical damage accumulation due to earthquakes
and fatigue and their coupling behaviors is necessary forthe appropriate maintenance. Therefore, CDM will be avalidmethod to evaluate the lifetime and thedeterioration of ultimate ability of concrete structures.
6.

ACKNOWLEDGMENT

The authors appreciate Penta-Ocean Construction Co., Ltd.
for corporation in testing the specimens, Nippon Steel
Composite Co., Ltd. for providing the CFsheet.
7.
[1]

8 cycles9 cycles11 cycles

15 cycles

Fig.17 Damage distribution(CASE1)
The relationship between cyclic fracture number andmaximum load is shown in Fig.15 and Fig.16.In both cases ,the analytical results are a good agreement comparable withthe experimental results. Finally, the damage distribution isshown in Fig.17.It was found that the adhesive fracture would
be evolved from the root (A part) to the tip (B part)
corresponding to cyclic number. In Fig.17, the white area
reveals the evolution of de-lamination or adhesive fracture.
Herein, the damage zone is enlarged to show up thedetachment from concrete surface.

5.

CONCLUSION

In this study, the 2-dimensional FEM implementing theelasto-plastic damage constitutive equation based on CDM
has been applied to the damage and fracture analysis of the
RC structural member and the block reinforced with CFsteet.
The following conclusions have been obtained:(1) The stress-strain relations of the concrete and the
CFsheet have been successfully identified by theelasto-plastic damage constitutive equation using
Drucker-Prager equivalent stress.
(2) The calculated damage zones have corresponded wellwith the experimental crack propagation in the RC slab
from which it is seen that the damage mechanics model
is valid for the damage evaluation of the concrete
structural member reinforced with the CFsheet.
(3) The adhesive fracture mode has been identified by ananother equivalent stress and which has corresponded
well with experimental fatigue fracture number. The
validity of present analysis has especially been
demonstrated by the fact that adhesive fracture has wellbeen simulated.
(4) The repair and the maintenance to increase the lifetime
are important technologies for concrete structures. The

REFERENCES

Liu,Yuqing.et al.,” Nonlinear Finite Element Analysis of Reinforced
Concrete Members Strengthened with Carbon Fiber Sheets,”
Proceedings of the Japan Concrete Institute,Vol.20,No.3,1998,pp.1-6.
[2] Murata,K.et al.,” Research of Mechanics Characteristics of RC BeamStrengthened with Carbon Fiber Sheets under Dynamic Load,”
Proceedings of the 47th Science Council of Japan Materials Research
Conference,2003,pp.215-216.
[3] Abe,T.et al.,” Research of Mechanics Characteristics of RC BeamStrengthened with CFS and its Rehabilitation Effect,” Proceedings of
the 53rd National Congress of Theoretical & Applied
Mechanics,2004,pp.215-216.
[4] Polak,M.A.,Vecchio,F.J.,” Nonlinear Analysis of Reinforced-Concrete
Shells,” Journal of Structural Engineering, ASCE Vol.119,No.12,
1993,pp.3439-3462.
[5] Lemaitre,J.,” A course on Damage Mechanics ,Second Edition,”
Springer,1990.
[6] Cipollina,A.et al.,” A Simplified Damage Mechanics Approach to
Nonlinear
AnalysisofFrames,”Computers
andStructures,Vol.54,No.6,1995,pp.1113-1126.
[7] Peng,X.,Meyer,C.,” A Continuum Damage Mechanics Model forConcrete Reinforced with Randomly Distributed Short Fibers,”
Computers and Structures,Vol.78,No.4,2000,pp.505-515.
[8] Al-Gadhig,A.H.,Baluch,M.H.,” Damage Model for Monotonic andFatigue Response of High Strength Concrete,” International J. of
Damage Mechanics,Vol.9,2000,pp.57-78.
[9] Toi,Y. et al.,”Element-Size Independent Elasto-Plastic Damage
Analysis of Framed Structures, ”Transaction of the Japan society of
Mechanical Engineering, Series A ,Vol.67,No.653,2001,pp.8-15.
[10] Toi, Y.,Lee,J.G.,” Element-Size Independent Elasto-Plastic Damage
Behaviors of Framed Structures, ”ICCSA2005,Lectures Notes in
Computational Science 3483,Springer,2005,pp.1055-1064.
[11] Tanaka H,Toi Y. et al.,” Damage and fracture analysis of brittle
structural elements reinforced with carbon fiber sheets ”JOURNAL OFENVIRONMENT AND ENGINEERING,Vol.3,No.1,2008,pp111-122[12] Hidenori TANAKA,Yutaka TOI,”Adhesive Failure Analysis of
Structural Elements Reinforced with Carbon Fiber Sheets,”Transactionof the Japan society of Mechanical Engineering, Series
A ,Vol.72,No.724,2006,pp.198-204.

322

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Development of Functional Carbon Nanotubes -Asphalt Composites
Tatsuo SHIRAKAWA, Akio TADA and Noriyasu OKAZAKI
Department of Civil & Environmental Engineering, Kitami Institute of Technology



ABSTRACT: The purpose of this study is to prepare asphalt
composites containing CNTs and to elucidate the characteristic
properties of CNTs-dispersed asphalt composites. In the present
study, such composites were prepared using several types of asphalt
emulsions as a binder. As a result, both nonionic and anionicemulsions kept dispersed state, when combined with as-prepared
CNTs. However, cationic emulsions failed to keep dispersed state.
But after several additional systematic experiments, it was foundthat cationic emulsion successfully retained an adequate dispersion
state, when as-prepared CNTs were treated with acid solution andthe following incremental addition of the pretreated CNTs. The
three types of CNTs-asphalt composites demonstrated higher
performances in both penetration and microwave-absorption tests
than asphalt composites including carbon black powder instead of
CNTs, when the mass percentage of all carbon materials was thesame.

Keywords: carbon nanotubes -asphalt composites, asphalt
emulsion, acid treatment, penetration test, absorption of
microwave
1. INTRODUCTION
In recent years, nanocarbon materials such as carbon
nanotubes (CNTs) and carbon nanofibers (CNFs) have been
attracting much global attention as potential fillers for
improving the mechanical and electrical characteristics of
cement mortars in many countries [1]. However, there has
not been much investigation into the application of
nanocarbon materials for asphalt pavement [2].
CNTs are well known for their superior performances inmechanical, thermal, and electrical performance [3]. A CNT
has a very high aspect ratio and so it probably acts as a bridge
between grains of sand, suggesting that CNTs-enriched
asphalt have the potential to reinforce asphalt pavement.
Moreover, CNTs are highly electro-conductive and therefore
CNT-enriched asphalt will surely absorb microwave andincrease in temperature. Often, the resulting temperature ishigh enough to melt solid asphalt into liquid asphalt that can
combine grains of sand and gravel. This novel form ofasphalt manipulation may provide a new way to pave roads
with asphalt.
CNTs are thought to be too expensive to use as
construction materials. Recently, however, a new production
process, which will possibly lower the cost of CNTs, has beenreported [4]. The process, which was named direct methane
reforming (DMR) process, can provide CNTs as a by-productof hydrogen production from natural gas using catalysts. In
this process catalyst materials remain in as-prepared CNTs
and so the purification is necessary to obtain CNTs of highpurity. Especially, CNTs for the uses of a lithium ion battery,electronic devices, etc. should be of higher purity. CNTs
have a variety of uses. For some uses, CNTs of not so highpurity will serve their purpose. A paving material is one
example. In this case CNTs are used as a filler of highmechanical strength due to their high aspect ratio.
The purpose of this study is to prepare new asphalt
composites containing CNTs and to elucidate thecharacteristic properties of CNTs dispersed asphalt
composites.
Since asphalt is not fluid at ordinary
temperatures, it does not act as a binder of aggregate.
Therefore, asphalt is usually heated at elevated temperatures
to lower the viscosity before its use. Another methods of
fluidizing without heating are the addition of a solvent toasphalt and the emulsification of asphalt with water using
surface-active agents. Asphalt emulsions, namely, colloidal
dispersions of asphalt particles in water have low viscosity
and so are easily mixed with aggregate even at ordinary
temperatures. Asphalt emulsions were made by emulsifying
high-viscosity asphalt into water with the aid of
surface-active agents. The mixture of asphalt-emulsion andaggregate became very strong, because the aggregate becamebound together with asphalt after water was removed off.
In order to disperse CNTs into a liquid, surface-active
agents are frequently used. Although surface-active agentsthat are used in asphalt emulsions are not common to those of
CNTs, the former agents could be act as the latter agents. If
so, asphalt emulsions, which are inexpensive, are suitablematerials for adding of CNTs. In this study, such composites
were prepared using several types of asphalt emulsions
instead of conventional asphalt which needs preheating.
2. EXPERIMENTAL
2.1 Materialsa) Asphalt emulsions as binders
As asphalt binders, cationic, anionic and nonionic asphalt
emulsions were used (Table 1) [5]. All asphalt emulsions
were provided by Nichireki Corporation. It is noteworthy
that the pH of every asphalt emulsion is different; cationicemulsion is acidic, nonionic emulsion is neutral and anionicemulsion is alkaline.
b) Carbon fillers
CNTs and carbon black are used as carbon fillers (Table 2)[6]. The CNTs were produce by the direct methane reforming
reaction using an iron catalyst [4].
An industrial carbon black sample was provided by Tokai
Carbon Corporation. Its arithmetical mean particle size (the323Table 1 Characteristics of asphalt emulsions
Emulsion type

Cationic

NonionicFor
permeation
Tuck coat
For mixing
MK-2 （ JIS
K 2208）For mixing
MN-1 （ JIS
K 2208）Viscosity(B type)
25°C
(cP)
50.2

1.93

Reductionof massduringevaporation
(wt%)
67.06

19.9
178.9

2.40
4.66

50.49
57.70

35.5

6.90

56.60
Fig. 1 Agglomeration on cationic emulsion surface

525.0

Anionic

pH

12.89

65.05
composites was evaluated by measuring a rise in the surface
temperature. This is because microwave absorbed by asample is converted into thermal energy to result in atemperature rise.
Every sample was irradiated withmicrowave of 500W for 10 seconds. The temperature
measurements were done with a radiation thermometer.

Table 2 Bulk density measurements of carbon fillers

No.
A
B
C
D
E
F

Carbon filler

As-prepared CNTs
As-prepared CNTs after
milling
Acid treated CNTs
(HCl)
Acid treated CNTs
(HNO3)
Carbon black producedby the DMR method
Industrial carbon blackBulk
density(g/cm3)

Ratio to thedensity ofsample A

0.02891.00

0.10483.63

0.15035.20

0.15175.25

0.10303.57

0.3380

11.7

3. RESULTS AND DISCUSSIONS
3.1 Dispersion of carbon fillers into asphalt emulsions
Table 3 summarizes the dispersibility of the carbon fillers
to different types of asphalt emulsions.
3.1.1 Effects of asphalt emulsions

primary particles) was 28nm [7]. Another carbon black
sample also was produced by the direct methane reforming
reaction using another iron catalyst. The as-prepared CNTs
were treated with dilute solution of hydrochloric acid (HCl)or nitric acid (HNO3) followed by filtering with suction.
2.2 Preparation of carbon filler-asphalt composites
CNTs or carbon black of 0.40g was added into an asphalt
emulsion of 40g and then mixed using a magnetic stirrer. The
resulting samples were dried in a dry oven maintained at 70°C
until constant weight was reached.
2.3 Penetration test
The penetration test (JIS K 2208) was adopted as astandard test of hardness evaluation of asphalt materials.
2.4 Absorption of electric magnetic wave
The microwave-absorption ability of the filler-emulsiona) Nonionic emulsion
The two types of carbon black samples became dispersedin the nonionic emulsion by stirring in a short time. And
as-prepared CNTs also became dispersed in the nonionicemulsion by stirring in over 5min.
b) Anionic emulsion
In the case of anionic emulsion that has higher viscosity
than nonionic emulsion, both CNTs and two types of carbon
black samples also dispersed, although the necessary time forstirring was longer than in the case of nonionic emulsion.
c) Cationic emulsion
In the case of the cationic emulsions, however, the degreeof dispersion depended on both the kind of a carbon-filler andthe type of an asphalt emulsion.
The two types of carbon black samples were dispersed inboth MK-2 emulsion (JIS K 2208), which is frequently used
for surface treating of wearing course and modified asphalt
emulsion for tuck coat layer.
On the other hand, it was difficult to disperse theas-prepared CNTs were not dispersed instantly; CNTs
particles stayed for a while in the upper zone of the MK-2
emulsion and modified-asphalt emulsion (Fig. 1).
This suggests that the wettability of the CNTs by thecationic asphalt emulsion is not good.

324

Scale: 5μm

A)

C)

E)

B)

D)

F)

Fig. 2 Effects of milling or acid treatment on as-prepared CNTs (SEM: 5,000 times)
3.1.2 Effects of CNTs-addition methodsa) MK-2 emulsion
Although carbon black samples added at once wasdispersed into MK-2 emulsion regardless of the quantity,
CNTs failed to disperse when a large quantity of CNTs wasadded at one time. However, the same quantity of CNTs wasmanaged to disperse when the CNTs were subdivided into
several parts and adding incrementally all the parts by
repeating the following step; a part of CNTs is added and thenstirred until well-dispersed state is achieved.
b) Modified asphalt emulsion
In the case of modified asphalt emulsion, as-prepared
CNTs hardly dispersed in spite of adopting the same addition
method as in the case of MK-2 emulsion, although carbon
black samples added at once were dispersed regardless of its
quantity.
3.2 Microscopic gathering states of carbon fillersa) The SEM images of acid-solution treated CNTs
Figure 2A shows an image of as-prepared CNTs and thisimage demonstrates that the bulk density of the CNTs is low.
Figure 2B shows an image of as-prepared CNTs that weremechanically milled prior to SEM examination. Figures 2C
and 2D are an image of the CNTs that were treated with dilutehydrochloric acid (HCl) and dilute nitric acid (HNO3)
respectively. As seen in Figs. 2C and 2D, the bulk density of
the acid-solution treated CNTs seems to have increased, andthere are a few tangles of the fibers. This situation seems to
be a good condition to scatter the CNTs in asphalt emulsion,
resulting in a good dispersion. Figures 2E and 2F show theimages of homemade carbon black and industrial carbon
black respectively. The shape of their samples is clearly
different from that of CNTs (seen in Figs. 2A-2D). The
former is spherical, while the latter is fibrous.
3.3 CNTs and their acid-solution treatmenta) SEM images of carbon fillers
Figure 2 shows a scanning electron microscopy (SEM)
image of as-prepared CNTs with a magnification of 20,000.
A secondary electron is discharged from iron particles
attached on tops of the CNTs and whitens with the SEM
image. Each iron particle was a constituent of the ironcatalysts used for producing the CNTs, clearly indicating thatcatalysts remain in the as-prepared CNTs.
A result of the analysis with the energy dispersion type
fluorescence X-ray analyzer (XRF-EDX) revealed that
as-prepared CNTs included a catalyst (iron: 4.3wt%).
In Fig. 2, the CNTs (diameter: ~100nm) flex in the shape ofa fiber and grow up while twisting and tangling. When the
CNTs twist up, a cavity occurs in the CNTs. As a result, thebulk density of the CNTs is low.
b) Acid–solution treatment of CNTs
As a result, the bulk density of the CNTs shrinks. Therefore,
it is thought that preprocessing to reduce the amount of325because the bulky agglomerate of as-prepared CNTs wasdisentangled and compacted.
3.4 Degree of dispersion of acid-solution treated CNTs inemulsionsa) Nonionic and anionic emulsions
The CNTs became dispersed in the nonionic emulsion afterstirring in a short time, indicating the affinity of both filler
with the liquid is very high. In the case of anionic emulsionthat has higher viscosity than nonionic emulsion, both CNTs
needed longer time than in nonionic emulsion to reach asimilarly state of dispersion.
b) Cationic Emulsions
In the case of the emulsion of MK-2 or tuck coats,
acid-solution treated CNTs fairly dispersed when a large
quantity of the CNTs was added at one time. And the degreeof dispersion became good when the CNTs were added
incrementally.
In the case of modified-asphalt emulsion, acid-solution
treated CNTs hardly dispersed when a large quantity of the
CNTs was added at one time. However, the CNTs fairly
dispersed when the CNTs were added incrementally.

Fig. 3 SEM image of as-prepared CNTs (20,000 times)

3.5 Improved dispersibility of acid-solution treated CNTs
into cationic emulsions
The acid solution dissolved iron particles in as-prepared
CNTs, releasing ferric ions (Fig. 4). In the same manner, thecationic emulsions, which are strong acid, probably dissolves
iron particles in as-prepared CNTs.
The ferric ions will interact with a surface-active agent that
contributed to stabilize the cationic emulsions and take away
its important role, resulting in the agglomeration. The reason
as-prepared CNTs failed to disperse is that the ferric ionsreleased from the residual catalyst in as-prepared CNTs.
This change in bulk density probably occurs because thebulky agglomerate of as-prepared CNTs was disentangled
and compacted.

Fig. 4 Treatment of as-prepared CNTs with a dilutehydrochloric acid solution
remaining catalyst is important. An acid–solution treatmentof the CNTs was carried out in anticipation of the
improvement in degree of their dispersion. When the
as-prepared CNTs contact the acid solution (Fig. 4), every
iron particle in the residual catalyst (fine particles of iron) is
dissolved into the solution.
Two specimens of the acid-solution treated CNTs samples
were measured by the XRF-EDX method and it was foundthat the concentration of residual catalyst (iron) was 2.0wt%
and 1.7wt% respectively. The concentrations are almost as inhalf as that of as-prepared CNTs, indicating that the
acid-solution treatment is useful to improve the purity of the
CNTs.
c) Bulk density
Table 2 lists the results of measuring the bulk density of thesame samples shown in Fig. 2. The bulk density ofas-prepared CNTs increased after milling or treating with theacid-solution. This change in density probably occurs

3.6 Characteristics of CNTs -asphalt compositesa) Penetration test
Figure 5 shows the results of penetration test (JIS K 2208)
for CNTs-emulsion composites.
Apparently a CNTsemulsion composite was more greatly stiffened by adding
CNTs than carbon black of the same mass concentration. In
other words, CNTs can give the same penetration degree ascarbon black samples with a lesser quantity of addition.
b) The absorption of microwave by CNTs-emulsion
composites
The experimental results are shown in Fig. 6 in whichsurface temperature was plotted as a function of time on asemi logarithmic graph. Except the anionic emulsion, thesurface temperature of CNTs-emulsion composites increased
with time, indicating that microwave was absorbed by thosecomposites.
3.7 Strong points of CNTs-asphalt emulsions
The above-mentioned results clearly show that CNTs arepotential filler for reinforcing as well as for giving

326

Table 3 Dispersibility of carbon fillers in asphalt emulsions
As-prepared CNTs

Tuck coat

2.40

Poor

For mixing
4.66
MK-2（JIS K 2208）
For mixing
6.90
MN-1（JIS K 2208）
12.89

Poor

Good

Good

Fair

Fair

Good

Good

Excellent--

Excellent--

Excellent

Good

--

Good

--

Good

CNTs 1.0wt%
Carbon black 1.0wt%
Control

Penetration degree （1/10mm）

140
120
100
80
60
40
20

Cationic

For mixing
MN-1

For mixing
MK-2

Tuck coatForpermeation0

Nonionic Anionic1000
CNTs 1.0wt%
Carbon black 1.0wt%
Control

100

10

1

0.1

Cationic

Fig. 5 Results of penetration tests

For mixing
MN-1

180
160

Industrial
carbon black
Addition at
once
Fair

Fair

For mixing
MK-2

Anionic

1.93

Incremental
addition
Poor(Agglomerate)
Fair

For
permeation

Nonionic

For permeation

Additionat once
Poor

Surface temperature change after microirradiation ΔT （K）

Cationic

pH

Tuck coat

Emulsion type

Acid treated CNTs (withHCl or HNO3)
Addition
Incremental
at once
addition
Poor
Fair

Nonionic AnionicFig. 6 Temperature change of carbon fillers by microwave
irradiation

microwave-absorbing property to conventional asphalt. Thisis probably the first study about CNTs-asphalt emulsions.
Asphalt emulsions, of which viscosity can be easily
controlled by the addition of surfactants, are a suitable binder
for asphalt pavement.
Civil structures of many countries need to renew old
facilities. To renew of civil structures, it has been proposedthat the implementation of new paving materials and methods
ought to be adopted for performance enhancement and life
cycle extension. The present study suggests that microwave
heating to lower asphalt viscosity for partial repairs may be a
promising construction technique. This will contribute to theease of pavement construction, improvements in safety,
and asphalt heating energy reduction. Thus, CNTs-asphalt
emulsions will be greatly useful for construction fields.

4. CONCLUSIONS
The main conclusion to be drawn from the above results is
as follows.
(1) In order to take advantage of the essential functions of
the CNTs involved in asphalt, the CNTs should be
dispersed into asphalt emulsions.
(2) As for both nonionic and anionic asphalt emulsions,as-prepared CNTs easily dispersed.
For cationicemulsions, however, as-prepared CNTs failed to disperse.
This problem was managed to solve by pretreating the
CNTs with the acid solution and adding incrementallythe pretreated CNTs.
(3) The CNTs-asphalt composites showed not only higher
stiffness (higher performance in the penetration test) thancarbon black-asphalt composites but also a new function,327

namely, the ability of microwave absorption, whencompared under the same conditions (mass concentrationof the each carbon material, measuring method, etc.).

5. ACKNOWLEDGMENT
A part of this study was supported by Grants-in-Aid forYoung Scientists (B), JSPS (No.23760401).

6. REFERENCES
[1]
[2]

[3]

[4]
[5]
[6]
[7]

MANZUR T, YAZDANI N. “Strength Enhancement of Cement
Mortar with Carbon Nanotubes Early Results and Potential”
Transportation Research Record, No.2142, 2010, pp.102-108.
Shirakawa T, Tada A, Okazaki N, Iwahashi T, Kohata N.
“Development of Construction Materials Using Advanced Nanocarbon
Provided in a Hydrogen Manufacturing Process Subsidiary” Journal of
JSCE Global Environment Engineering Research, Vol. 18, Aug.2010,
pp. 81-88.(in Japanese)
Wu Z, Chen1 Z, Du X, Logan J M, Sippel J , Nikolou M, Kamaras K,
Reynolds J R, Tanner D B, Hebard A F, Rinzler A G. “Transparent,
conductive carbon nanotubes films”, Science, Vol.305, Aug. 2004,
pp.1273-1276.
Tada A, Matsunaga T, Okazaki N. “Direct methane reforming process
and its applications”, Transactions of the Materials Research Society ofJapan, Vol.33, No.4, Dec. 2008, pp. 1059-1062.
AEMA, MS-19 Basic Asphalt Emulsion Manual, Asphalt Institute,
2008.
Gridley P F, Vallerga B A. “Carbon black reinforcement of asphalts inpaving mixtures”, ASTM Special Technical Publications, No.724,
1980, pp.110-128.
Yamaguchi K, Sasaki I, Nishizaki I, Meiarashi S, Moriyoshi A.
“Effects of film thickness, wave-length, and carbon black on photo
degradation of asphalt”, Journal of the Japan Petroleum Institute,
Vol.48, No.3, 2005, pp.150-155.

328

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Hydraulic Character of Estimation Method on Roughness Coefficient ofConcrete Canal

KOICHIRO OTAGAKI, EHIME University, United graduate school of agricultural sciences
KOJI SAI, KOCHI University, Faculty of agriculture
KENICHI FUJISAWA, NIPPON JIKKOU Co. Ltd
SHINSUKE MATSUMOTO, SHUSHI SATO, KOCHI University, Faculty of agriculture

condition of canal [3].
According to some previous studies [4], it was reported
that the values of roughness coefficient varied due tohydraulic conductive conditions of canal. To establish
evaluation method of hydraulic performance, it is importantto verify the relationships among roughness coefficient,
hydraulic condition and canal surface condition.
This study aimed to establish the prescribed method on
hydraulic performance of agricultural concrete canal.
Through the various experiment under several hydraulicconditions,roughnesscoefficientwasestimated
quantitatively. Moreover, some of the subjects on theestimation of this coefficient were extracted.

ABSTRACT: The agricultural concrete canals are approximately 450

thousand km in Japan. Recently, aging and deformation of them is
becoming a subject to solve, it is necessary to preserve theirfunctions appropriately. Estimation method of hydraulicperformance of concrete canal by roughness coefficient was aimed
in this research. Hydraulic roughness coefficients of concrete canalwere estimated under several hydraulic conditions in theexperiment. Roughness coefficient was calculated using measured
data of water surface slope. Reynolds number and Froude number
were particularly focused in this study. As a result, roughness
coefficient was varied enormously with the differences of hydraulicconditions. Especially, Reynolds number and Froude number
represent the characteristics of water flow had strong negative
correlation with roughness coefficient. It was also suggested thatthe shape transformation at the downstream of the canal affected toroughness coefficient.

2. METHOD AND PROCEDURE
Keywords: Concrete canal, Hydraulic performance, Flowtransmissibility, Roughness coefficient, Stock management
1. INTRODUCTION
Agricultural concrete canals are important infrastructure
for stable food production. In Japanese agricultural area,
aging and drops of function of canals were observed as recent
problems, establishment of appropriate maintenance andmanagement scheme is required for the preservation of thesefacilities. To solve this subject, it is necessary to evaluate
quantitatively its present performance as well as to choose theoptimum countermeasures against the factor of performance
declination.
Main performances of agricultural concrete canal are
Hydraulic, Water utilization, Structural safety, and Social
safety. Particularly, hydraulic performance is very importantas it is affected to the water supply to farmland constantly.
In general, roughness coefficient in Manning mean
velocity formula was used as an index of hydraulicperformance. Roughness coefficient is well known as abarometer of water conduction. On the other hand, Manningequation is used in canal design and roughness coefficients ofvarious materials are defined in public guideline [1], [2].
Roughness coefficient was also used for evaluation of surface

2.1 Outline of experiment
For the estimation of roughness coefficient, the hydraulicexperiments were conducted using slope variable canalsystem. In the experiment, two patterns of canal surface
conditions were examined; one is the surface covered withpolyurethane coating and acrylic board and the other one is
the surface of concrete flume. The roughness coefficient wascalculated using observed data of water surface profiled ateach surface condition. The outline of experiment is shown in
Fig. 1. Hydraulic conditions were set up by changing crosssectional average velocity and canal slope under the conditionof constant water quantity. Water surface profiles wereobserved at intervals of 1m by point gauge (KENEK Co.
PH-340). The observed points of water surface profile of P0 to
P7 were shown in Fig.1. In the each measurement point of P0
to P7, mean depth were obtained by calculation the average ofthe three points in cross direction that shown in Fig.2.
2.2 Calculation method of roughness coefficient
In this study, the roughness coefficient was estimated bysolving fundamental equation of non-uniform flow using
observed data of water surface profiles. The fundamental
equation of non-uniform flow is described as follow;329

where, I 0 is canal bottom slope, I w is water surface slope
which is average of gradients of P0 to P7 in the Fig.1and Fr is
Froude number described as (  v gh )．
3 RESULTS AND DISCUSSIONS

Fig. 1 Overview of experimental channel andmeasurement points of water depth.

Fig. 2 Measurement points of water depth in transverse
section.
i
dh

dx

n 2Q 2

bh /( b  2 h )4 3 b 2 h 2
1

Q 2

(1)

2 3

gb h

where x is the direction of downward，h is water depth(m)，Q is water quantity(m3/s)，α is flow coefficient(=1)，g isgravitational acceleration(m/s2)，b is canal breadth(m)，ｉ iscanal bottom slope，and n is roughness coefficient(s/m1/3).
The estimation procedure of roughness coefficient is asfollows; using fluctuated value of n, water depth hc were
calculated by integrating the Equation(1) in x=1(m).
Evaluation function concerning observed water depth h0 andcalculated water depth hc. The evaluation function is
described as follows;E

1 N h0,i  hc,i
N i 0
h0,i



2

(2)

where，N is number of data, h0 is observed water depth(m),hc is calculated water depth(m) andｉ is a number of data.
Thus, the value of n is changed, and the above operation isrepeated. Finally, the value of n which outputs the smallest
evaluation function is defined as appropriate values ofroughness coefficient.
On the other hand, convenient method for the estimationof roughness coefficient of actual agricultural concrete canalis, in general, solving Manning’s formula [5]. Therefore,roughness coefficient was also estimated from Manning’s
formula by assumption of water flow in the experiment was
uniform flow. Energy gradient in the formula was calculated
as follows;I  I w  Fr 2 I 0  I w (3)

3.1 Result of estimated roughness coefficient.
Estimated roughness coefficient obtained from experiment
under several canal surface conditions of polyurethane
coating and concrete flume were shown in Table 1 and 2,
respectively. In these tables, n1 is calculated by thefundamental equation of non-uniform flow, and n2 iscalculated by Manning’s formula. It was difficult to observe
when Froude number is 0.4 or higher because of unstablewater surface. Therefore, all the latter results are discussed
when the Floude number was smaller than 0.4.
As a result, the estimated roughness coefficient changed
greatly under the different hydraulic conditions even though
the surface condition of canal was same. Results of roughness
coefficient shown in Table 1 and 2 clarified that large values
comparing with the designed values were confirmed when
they were estimated under the most of all hydraulicconditions.
3.2 Estimation method of roughness coefficientComparison result on estimation methods of roughness
coefficient were examined based on the Figure 3. In thisfigure, the relations between n1 and n2 (  n 2  n1 ) and water
surface slope were appeared. The results show that thedifferences between n1 and n2 increased as the slope of water
surface become steep. These results suggested that estimation
method by Manning’s formula might be uncertain under thecondition of non-uniform flow. Therefore, it needs to pay
attention to hydraulic condition and form and slope of thecanals when using this equation.
3.3 Influence of hydraulic conditions to estimatedroughness coefficient.
In the same condition of canal bottom slope, estimatedroughness coefficient changed as water velocity was varied.
This result suggests that flow condition of canal has affectedenormously to the estimation of roughness coefficients.
Figure 4 shows the relation between hydraulic mean radiusand roughness coefficient. There is a high correlation
between hydraulic mean radius and roughness coefficient. On
the other hand, it wasn’t confirmed the large difference
between polyurethane coating and concrete flume. As aresult, roughness coefficient is affected by hydraulicconditions extremely. Moreover, it was suggested that
estimated roughness coefficient could not express the exact
loss of water flow due to surface condition.
3.4 Influence of hydraulic parameters to roughness
coefficient
To examine with considering the mechanical resemblance offlow, relationships between roughness coefficient andnon-dimensional hydraulic parameter, Floude number Fr andReynolods number Re , were evaluated.

330

Table 1 Hydraulic conditions and estimated values of roughness coefficientin case of experiment using experimental channel.
Experimental
conditions

Channel slope（×10-3）Water
velocity(m/s)

Meandepth(m)Hydraulicradius(m)

0.00
0.00
0.00
1.75
1.75
1.75
3.49
3.49
3.49
5.24
5.24
5.24

0.080
0.099
0.126
0.086
0.109
0.152
0.090
0.119
0.185
0.098
0.133
0.257

0.222
0.181
0.127
0.208
0.165
0.112
0.198
0.150
0.097
0.183
0.134
0.070

0.150
0.130
0.100
0.143
0.121
0.090
0.138
0.113
0.080
0.131
0.104
0.061

(1)
(2)
(3)
(4)
(5)
(6)
(7)
(8)
(9)
(10)
(11)
(12)

Froudenumber

0.05
0.07
0.13
0.06
0.09
0.15
0.07
0.10
0.19
0.07
0.12
0.33

Reynolds
number

10288
10958
11962
10511
11247
12278
10673
11521
12628
10926
11829
13290

n1

n2

0.07640.05740.03590.07290.05220.03070.0676
0.04680.02040.05970.03790.00950.11250.07960.04400.16200.10670.05720.18130.11440.0438

Table 2 Hydraulic conditions and estimated values of roughness coefficientin case of experiment using concrete bench flume.
Experimental
conditions

Channel slope（×10-3）Water
velocity(m/s)

Meandepth(m)

0.00
0.00
0.00
0.00
1.75
1.75
1.75
1.75
3.49
3.49
3.49
3.49
5.24
5.24
5.24

0.111
0.140
0.189
0.202
0.118
0.151
0.185
0.244
0.126
0.164
0.236
0.297
0.137
0.183
0.286

0.247
0.197
0.146
0.123
0.232
0.181
0.131
0.109
0.217
0.167
0.116
0.092
0.200
0.150
0.096

The difference between n2 and n1

(1)
(2)
(3)
(4)
(5)
(6)
(7)
(8)
(9)
(10)
(11)
(12)
(13)
(14)
(15)

Hydraulicradius(m)

0.135
0.119
0.098
0.087
0.131
0.113
0.091
0.080
0.126
0.107
0.084
0.070
0.120
0.100
0.073

: Experimental results of concrete bench flume: Experimental results of experimental channel
0.15

0.10

0.05

0.00
–6

–5

–4
–3
Surface slope

–2

–1
0
–3
（×10 ）

Fig. 3 Differences between n1 and n2 under conditionsof each water inclination.
Reynolds number is described as follows;
Re 

vR



(4)

Froudenumber

0.07
0.10
0.16
0.20
0.08
0.11
0.19
0.24
0.09
0.13
0.22
0.32
0.10
0.15
0.31

Reynolds
number

12851
14145
15765
16586
13194
14589
16308
17183
13585
15032
16870
17908
14037
15621
17757

n1

n2

0.05760.04220.02960.02520.05750.04230.02580.02240.05120.03560.02320.01840.04640.03060.01370.07580.05210.03180.02350.10820.07530.04340.03000.12150.08070.0431

where  is kinematic viscosity coefficient(m2/s). Figure 5
shows the relationship between Froude number androughness coefficient, Figure 6 shows the relation between
Reynolds number and roughness coefficient. From the resultsof Figure 5 and 6, both Froude number and Reynolds number
has highly negative correlation with roughness coefficient.
Correlation coefficient between Froude number androughness coefficient on concrete flume is 0.93, and on thepolyurethane coating is 0.90. Correlation coefficient between
Reynolds number and roughness coefficient on concrete
flume is 0.99, furthermore, on polyurethane coating is 1.00.
From the results of these experiments, extremely large
roughness coefficients were observed when Froude number
was within the small value. These results may cause by
influence of back water or drop down of water due to theshape change at the end of canal. Hydraulic conditions werechanged by heading-up gate at the end of the canal. However,equation (1) is representing the water slope caused of surface
roughness friction; the variation of water surface due to the331For establishment of estimation method of roughness
coefficient, an appropriate experimental condition must be
investigated in the future.

: Experimental results of concrete bench flume: Experimental results of experimental channel
Roughness coefficient n1

0.08
0.07
0.06

4 CONCLUSIONAs the fundamental study for achieving the verification of
hydraulic performance on agricultural concrete canals, theinfluence of hydraulic conditions and parameters toroughness coefficient were studied quantitatively. The resultsof experiments are concluded as follows;(1) Using the convenient estimation method of roughness
coefficient according to Manning’s formula, the difference
between measured and estimated roughness coefficientincreased as the increment of water surface gradient.
However, this simple equation is available for the practicaluse. Therefore, it is necessary to define the appropriate
hydraulic conditions for using this equation.
(2) Estimated roughness coefficients had varied enormously
as the differences of hydraulic conditions such as velocity andhydraulic mean radius.
(3) Roughness coefficient value had strong relationship with
Froude number and Reynolds number.
(4) In the case of estimating roughness coefficient on
experiment canal in finite length, roughness coefficient may
be estimated at excessive values due to the change of the form
at the end of canal by weir board.
From this study, the roughness coefficient of canals changed
widely due to hydraulic condition. Hereafter, from theviewpoint of hydraulic, it is absolutely necessary to find out
characteristic of changing the value of roughness coefficientand appropriate hydraulic conditions for estimation.
Furthermore, it is important to accumulate data of roughness
coefficient obtained several surface conditions of canal toestablish estimating method of roughness coefficient.

0.05
0.04
0.03
0.02
0.01
0.06 0.07 0.08 0.09 0.1 0.11 0.12 0.13 0.14 0.15
Hydraulic radius（m）

Fig. 4 Relationship between hydraulic radius androughness coefficient.

: Experimental results of concrete bench flume: Experimental results of experimental channel
Roughness coefficient n1

0.08
0.07
0.06
0.05
0.04
0.03
0.02
0.01
0.1

0.2
Froude number

0.3–3（×10 ）

Fig. 5 Relationship between Froude number androughness coefficient.
: Experimental results of concrete bench flume: Experimental results of experimental channel
Roughness coefficient n1

0.08
0.07
0.06
0.05
0.04

5 ACKNOWLEDGMENTS

0.03

A part of this study was supported by JSPS KAKENHI(23380143) (Grant-in-Aid for Scientific Research (B), A
representative; Professor S. Matsumoto). Also a part of thispaper is based on the discretionary spending from president ofKochi University in 2008 and 2009, and discretionary
spending from president of faculty of agriculture of Kochi
University in 2010.

0.02
0.01
10000

12000

14000
16000
Reynold number

18000

20000

Fig. 6 Relationship between Reynolds number androughness coefficient.

formation changes of cross-section of canal was notconsidered well in the experimental data. Therefore, apparent
roughness coefficient increased while the flowing spot of low
Froude number where the flow was strongly affected by thelower stream. That is why the roughness coefficient cannot
be estimated accurately. Further examinations concerned to
the variation characteristic of estimated roughness coefficient
will be necessary from the viewpoint of various hydraulicconditions.
It is very difficult to exclude the influence of downstream on
canal experiment. However, Figure 5 shows that hydraulicconditions of the range over 0.3 of Froude number conduct
the values of roughness coefficient become stable.

6 REFERENCES
[1]
[2]

[3]
[4]

[5]

332

JAPAN SOCIETY OF CIVIL ENGINEERS, “The collection of
hydraulic formulas”, Maruzen, 1986.
“Design guideline of Land improvement projects federation, canaldesign”, Ministry of Agriculture, Forestry and Fisheries society ofirrigation, Drainage and rural engineering ,2001, pp.152-207
HORACE W. KING and ERRNEST F. BRATER “Handbook of
Hydraulics, 5th”, McGraw-Hill Book Co. Inc., 1963, p.7-18.
TAKASHI KATO, SHINYA HONMA, KITAMURA KOJI and
IMAIZUMI Masayuki, “Deterioration of Irrigation Canals and Changein Roughness Coefficient”, Report of the NIRE, No.275, 2008, pp.
183-193.
TAKESHI TAKEMURA, HAJIME TANJI, YOSHINOBU ARARAGI,“Study on a method to estimate coefficients of roughness,” Journal ofJapanese society of irrigation, DRAINAGE AND RURALENGINEERING, 2001, pp. 25-28.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Na-bentonite as Rock Containment Barrier against Heavy Metals andAcid Leachates
Angelica Naka1, Takeshi Katsumi2, Toru Inui3, Giancarlo Flores4, and Takehiro Ohta5



ABSTRACT: Contamination by acid rock drainage (ARD) is
recognized as one of the most serious environmental problems
facing mining and some specialized construction industries, such as
in the case of road and rail transport. These activities involve
excavation and transport of large amount of rocks and sediments, so
if they contain heavy metals of natural origin, a proper disposal isnecessary to prevent the oxidation of sulfide minerals, and thesubsequent release of acid leachates with high metal concentrations.
The use of Na-bentonite as rock containment barrier seems to be apossible solution since previous studies have found that it has large
specific surface area and low hydraulic conductivity. However,since the long-term performance of this mineral when exposed to
ARDs is not widely understood, the aim of this research is toinvestigate its long-term barrier performance against these acidleachates. For this purpose chemical and physical parameters, suchas metal attenuation capacity and hydraulic conductivity on naturaland artificial ARD were evaluated.

Keywords: acid rock drainage, geosyntehtic clay liner,
bentonite, metal sorption, hydraulic conductivity
1. INTRODUCTION
The need to control natural contamination by metals andmetalloids coming from mine tailings and rock disposal sites
represents a current challenge for the mining industry andconstruction companies in many countries such as Japan [1].
In the past, remediation technologies were focused on
physical covers to reduce the production of acids by limiting
infiltration of water and oxygen [2]. However, recent
researches have suggested that potentially toxic elements,
particularly As, Se, and in some cases Ni, and Zn are mobile
even under neutral pH-conditions [2], [3]. Moreover, thereductive dissolution of As-bearing minerals has resulted in
the release of As [3].
As a result, disposal of excavated rocks with potential ofacid rock drainage (ARD) from construction and mining ismoving towards storage of hazardous materials in lined
containment facilities [4]. To this effect, the use ofgeosynthetic clay liners (GCLs), which usually containNa-bentonite, seems to be a good solution, since previous
studies have suggested that Na-bentonite has a relatively
large specific surface area, high metal sorption capacity,1
Graduate Student, Graduate School of Global Environmental Studies, Kyoto
University, Japan
2
Professor, Graduate School of Global Environmental Studies, Kyoto
University, Japan
3
Associate Professor, Graduate School of Global Environmental Studies,
Kyoto University, Japan
4
Lecturer, Graduate School of Engineering, Kyoto University, Japan
5
Senior Researcher, Disaster Prevention Technology Division, Railway
Technical Research Institute, Japan

self-healing capacity, and low hydraulic conductivity, and isavailable in many parts of the world [5], [6], [7].
GCLs have been commonly used in landfills. Nowadays,
their potential use in mining operations is being tested [2], [4],[8] since it is not a simple matter of transferring thetechnology from landfill applications to rock containmentswith potential of ARD generation [9], due to the extreme
range in leachate properties. Therefore this research aims toinvestigate the barrier performance of Na-bentonite when
exposed to ARDs (natural and artificial) with different metalcomposition and pH values through batch sorption tests and9-month hydraulic conductivity tests.
2. MATERIALS
Na-bentonite was obtained from a needle-punchedgeosynthetic clay liner (Bentofix® NSP 4900), which is afiber-reinforced GCL. This GCL contained powered sodium
bentonite sandwiched between woven and non-woven
geotextiles, and had a unit mass of 4670 g bentonite/m2.
Bentonite within this GCL had a water content, specificgravity, and smectite content of approximately 10.0%, 2.85,
and 80%, respectively.
Nine different natural rock leachates obtained fromdifferent parts of Japan were used. Four of them were liquidsamples, collected from four ore deposits (Kaminosawa,Okunosawa, Honko and Tateishi) of Kamikita mining
complex (Aomori Prefecture) on July 2010. The other fiverock drainage samples were obtained after leaching test using
rocks from excavated sites and natural ground located in
Yamanashi, Hyogo, Miyagi (2 samples) and Tokyo. The
rocks from excavated sites and natural ground were crushed
under 2 mm diameter and dried in the oven at 105±5 °C for 1day, and 150 grams of these material were mixed with 1.5 Lof distilled water in a 2 L plastic bottle with hermetic cap andplaced in an incubator shaker at 100 rpm for 40 days at 20°C.
The pH, EC, and ORP were monitored periodically. Before
conducting sorption tests, the mixtures were centrifuged andfiltrated, and the metal contents were analyzed by ICP-MS(Agilent 7500ce). Some chemical properties of the naturaldrainage from mines and excavated rocks and natural groundfrom different sites of Japan are summarized in Table 1.
These values were considered as the initial conditions.
An artificial ARD was prepared in the laboratory based on
the drainage composition of a Pb-Zn-(Cu) deposit located in
Cerro de Pasco, Peru [10]. Metal composition is presented inTable 2. GR grade FeSO4·7H2O, Al2(SO4)3·16H2O,
CuSO4·5H2O, ZnSO4·7H2O, Na2HAsO4·7H2O, PbCl2, K2SO4,
Na2SO4, CaSO4, and MgSO4 were mixed, and then the pH
was adjusted to 3 using H2SO4. EC was 1195 mS/m.

333

(b) Ikuno(a) Mokura
1x10

1x10 1

Metal concentration (mmol/L)

Metal concentration (mmol/L)

1x10 10

1x10 -11x10-21x10 -3
1x10

-4

1x10

-5

Before
After
Na

Mg

Al

K

Ca

Fe

Cu

Zn

As

1x10 01x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5

Pb

Na

Mg

Al

K

Element

1x10

Metal concentration (mmol/L)

Metal concentration (mmol/L)

Zn

As

Pb

1x10 10

1x10 -11x10-21x10 -31x10-4BeforeAfter1x10 -5
Na

Mg

Al

K

Ca

Fe

Cu

Zn

As

1x10 01x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5

Pb

Na

Mg

Al

K

Element

Metal concentration (mmol/L)-1

1x10 -2-3

1x10 -4
1x10

Before
After

-5

Na

Zn

As

Pb

Mg

Al

K

Ca

Fe

Cu

Zn

As

1x10 01x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5

Pb

Na

Mg

Al

K

Element

Metal concentration (mmol/L)

1x10 -1
1x10
1x10

-3

1x10 -4

Before
After

-5

Na

Zn

As

Pb

Mg

Al

K

Ca

Fe

Cu

Zn

As

Cu

Zn

As

Pb

1x10 01x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5

Pb

Na

Mg

Al

K

ElementCa

Fe

Element(i) Tateishi(j) Artificial ARD

1x10 11x10 2

Metal concentration (mmol/L)

1x10

Cu

1x10 10-2Fe(h) Honko

1x10 1
1x10

CaElement(g) Okunosawa

1x10

Cu

1x10 11x10 0

1x10Fe(f)Kamisosawa

1x10 1

1x10

CaElement(e) Aobayama
Metal concentration (mmol/L)Cu(d) Shintera

1x10 1

Metal concentration (mmol/L)Fe

Element(c) Core

Metal concentration (mmol/L)Ca

0

1x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5
Na

Mg

Al

K

Ca

Fe

Cu

Zn

As

1x10 11x10 01x10 -1
1x10 -2
1x10 -3
1x10 -4

BeforeAfter1x10 -5Pb

Na

Element

Mg

Al

K

Ca

Fe

Cu

Zn

As

Pb

Element

Fig. 1. Metals sorption on bentonite: (a) Mokura, (b) Ikuno, (c) Core, (d) Shintera, (e) Aobayama, (f) Kaminosawa,(g) Okunosawa, (h) Honko, (i) Tateishi, (j) Artificial ARD
334

Table 1. Profile and chemical characteristics of natural ARDs from Japan
Identification
Kaminosawa
Okunosawa
Honko
Tateishi
MokuraIkuno
Aobayama
Shintera
Core

Place
Aomori
Aomori
Aomori
Aomori
Yamanashi
Hyogo
Miyagi
MiyagiTokyo

Type
LiquidLiquidLiquidLiquidAltered volcanic rock
Altered volcanic rock
Mud sediment
Mud sediment
Mud sediment

pH
2.37
1.23
2.55
1.80
7.70
5.16
3.33
8.09
6.25

EC (mS/m)
297
1227
208
620
9.52
57.2
213
51.6
182.1

ORP (mV)
10
81
48
69
156
262
354
140
205

3. SORPTION TEST

4 HYDRAULIC CONDUCTIVITY TESTThe chemical performance of bentonite was studied
through batch sorption tests. Sorption test is a quick methodthat provides information about bentonite performance
regarding metal affinity, metal sorption capacity and involved
mechanism. For this test, 10 different ARDs (9 natural and 1
artificial ARD) were used.
Sorption tests were conducted using 0.1 g of bentonite in
50 mL of solution for the natural ARDs and 0.2 g bentonite in
50 mL of solution for the artificial ARD (higher amount in thelatter case because of higher metal concentration). Samples
were taken after 24 hours on an incubator shaker at 100 rpm
and 25ºC. After shaking, every mixture was centrifuged andfiltered using a filter with a 0.22-µm pore size. The
concentrations of Fe, Cu. Zn, Al, As, Pb, Na, Ca, Mg, and K
before and after the sorption tests were analyzed by ICP-MS(Agilent 7500ce).
The reduction in the concentration values of 10 metals
after the addition of bentonite is presented in Fig. 1. In somecases, the reduction was small, which suggests that thesolid-liquid ratio was low, but enough to detect sorption.
Results led to a better understanding of sorption capacity ofbentonite against heavy metals in a metal complex system(natural leachates from excavated rocks and mines andartificial ARD). Due to the release of Na, K, Ca, and Mg,
cations present in Na-bentonite, it can be inferred that themechanism involved is probably ion exchange.
Sorption test results usually represent ideal conditions and,
therefore, to evaluate field application of bentonite, additional
tests, such as hydraulic conductivity test, become necessary.

Nine-month hydraulic conductivity tests were conductedon GCL following the procedures described in ASTM D 5084
“Standard Test Methods for Measurement of Hydraulic
Conductivity of Saturated Porous Materials Using a Flexible
Wall Permeameter” and ASTM D 7100 “Standard Test
Method for Hydraulic Conductivity Compatibility Testing ofSoils with Aqueous Solutions”, using a falling
headwater–constant tailwater system. Fig. 2 shows a typical
diagram for this system, which consists of a flexible-wallpermeameter with a cell pressure of 30 kPa and an average
hydraulic gradient of 85–95 at constant room temperature of
25°C. The GCL (specimen) was placed between filter papers,
geotextiles, and plastic caps and was confined by a latex
membrane on the sides. The thickness of the GCL wasmeasured regularly using a cathetometer, while the EC, pH,effluent volume, and metal content by ICP were measured
periodically (results of effluents not shown).
Table 3. Testing conditions for hydraulic conductivity test
SampleSample 1
Sample 2
Sample 3

Prehydration solution
ARD
Distilled waterDistilled water
Vent Port

O-Ring seal

Water

Membrane

Table 2. Artificial ARD composition

Influent

Metal
Al
Fe
Cu
Zn
As
Pb
K
Na
Ca
Mg

Concentration (mg/L)
259.2
4330.2
86.9
493.1
49.1
2.9
31.8
413.9
397.0
214.0

Permeant solution
ARD
ARD
Distilled waterSpecimen

Acrylic cylinder
Geotextile andFilter paper
Air pressure
Regulator

Water
Gauge
Effluent

Cell
pressureFig 2. Scheme of a flexible-wall permeameter
For hydraulic conductivity tests, the artificial ARD wasused because its composition is critical in terms of metal
concentration and pH values compared to the natural ARDs.
335

Because waste rock containment is expected to last long,
hydraulic conductivity, pH and EC changes, and metal
release over time were monitored for nine months. Before
ARD permeation, two types of prehydration were performed
using either ARD (Sample 1) or distilled water (Sample 2).
For the prehydration process, GCLs were placed into
containments with the corresponding permeant solution forseven days and 30 kPa of pressure was applies. As a control, atest with distilled water prehydrated and permeated GCL wasalso conducted (Sample 3).
The results of the hydraulic conductivity test are presented
in Fig. 3. The hydraulic conductivity of GCL permeated with
distilled water (control) was constant, with an average of
1.4×10-11 m/s. The hydraulic conductivity value of the GCL
prehydrated with water and permeated with ARD was around
1.1×10-10 m/s, and the hydraulic conductivity of the GCL
prehydrated and permeated with artificial ARD, around
5.0×10-10 m/s. The hydraulic conductivity of GCL
prehydrated with water and permeated with ARD was fivetimes lower than the one of ARD prehydrated and permeated
GCL, which suggests that water prehydration positively
impacts on hydraulic conductivity.
Hydraulic conductivity values of GCLs permeated with thenatural ARDs have not been determined yet, but can beprobably predicted, considering that hydraulic conductivity
values may be a function of EC, pH, or swell volume. For
example, according to a relationship between hydraulicconductivity and swell volume presented in [7], at a 8.5 mL/2
g bentonite of swell volume obtained for the artificial ARD(results presented in [6]), the expected hydraulic conductivity
value should be 3.6x10-10 m/s. This value is close to theexperimental value (5.0x10-10 m/s) with a 72% accuracy andconfirms the relationship between the hydraulic conductivity
and free swell index reported in the literature. Relationships
between parameters are still under investigation and, thus
constitute part of the future research.

Hydraulic conductivity (m/s)

1.0 x 10-8

Sample 1: prehydration (ARD) / permeation (ARD)
Sample 2: prehydration (DW) / permeation (ARD)
Sample 3: prehydration (DW) / permeation (DW)

1.0 x 10-9

1.0 x 10-10

1.0 x 10-11

1.0 x 10-12
0

200

400

600

800 (Sample 1)

0

50

100

150

200 (Sample 2)

10

20

30

40 (Sample 3)0

Pore volumes of flow (PVF)

Fig. 3. Hydraulic conductivity change of each condition

4 CONCLUSIONS AND FUTURE RESEARCH
From natural rock leachates and mine drainage it wasobserved that these minerals have the capacity to sorb heavy

metals even when they are in combination (complex metal
mixture).
Nine-month hydraulic conductivity tests showed thatpermeability changes over time, especially due toprecipitation (physical clogging). The hydraulic conductivity
remained low during the test duration and was approximately
5 times lower when GCL was prehydrated with water before
ARD permeation (1.1x10-10 m/s) than the case in which
prehydration and permeation were done using ARD(5.0x10-10 m/s).
Considering that bentonite inside GCLs has the potential toretain heavy metals present in solution, showing relatively
low hydraulic conductivity under even extreme conditions,with availability in many parts of the world, GCLs seem toprovide an alternative for barriers against rock leachates withhigh heavy metal content.
Future research will be done on the relationship between
parameters such as EC, pH, swell volume or metal sorption
capacity and hydraulic conductivity in order to predict field
application of GCLs.
5 ACKNOWLEDGMENTS
The Nikkei Scholarship granted by The Nippon Foundation(Nippon Zaidan) has been provided to the first author’s
postgraduate study. The authors express deep gratitude to Mr.
Atsushi Takai and Dr. Zhenze Li (Kyoto University), andTakuya Urakoshi and Tomokazu Ishihara (Railway TechnicalResearch Institute) for their helpful discussions andconstructive comments and suggestions.
6 REFERENCES[1] Ohta, T., Enomoto, H., and Tokunaga, T., 2006. Evaluation andprediction of pollution caused by acid water exuded from mud sediment inurban ground, IAEG 2006, No.265, pp.1-9.
[2] Lange, K., Rowe, R. K., and Jamieson, H., 2010. The potential role ofgeosynthetic clay liners in mine water treatment systems, Geotextiles andGeomembranes, Vol.28, No.2, pp.199-205.
[3] Rowe, R. K., 2006. Some factors affecting geosynthetics used forgeoenvironmental applications, 5th International Conference on
Environmental Geotechnics, London, pp.43-69.
[4] Lange, K., Rowe, R. K., and Jamieson, H., 2007. Metal retention ingeosynthetic clay liners following permeation by different mining solutions,
Geosynthetics International, Vol.14, No.3, pp.178-187.
[5] Shackelford, C. D., Sevick, G. W., and Eykholt, G. R., 2010. Hydraulicconductivity of geosynthetic clay liners to tailings impoundment solutions,Geotextiles and Geomembranes, Vol.28, No.2, pp.149-162.
[6] Naka, A., Li, Z., Inui, T., Katsumi, T., and Mogami, H., 2010. Heavy
metals retention in geosynthetic clay liners and its potential role in acid rockdrainage treatment, Geosynthetics Engineering Journal, Japan Chapter ofInternational Geosynthetics Society, Vol.25, pp.233-240.
[7] Katsumi, T., Ishimori, H., Ogawa, A., Yoshikawa, K., Hanamoto, K.,
and Fukagawa, R., 2007. Hydraulic conductivity of nonprehydrated
geosynthetic clay liners permeated with inorganic solutions and wasteleachates, Soils and Foundations, Vol.47, No.1, pp.79-96.
[8] Lange, K., Rowe, R. K., and Jamieson, H., 2009. Diffusion of metals ingeosynthetic clay liners, Geosynthetics International, Vol.16, No.1,
pp.11-27.
[9] Hornsey, W. P., Scheirs, J., Gates, W. P., and Bouazza, A., 2010. The
impact of mining solutions/liquors on geosynthetics, Geotextiles andGeomembranes, Vol.28, No.2, pp.191-198.
[10] Wibkirchen, C., Dold, B., Friese, K., and Glaber, W., 2005.
Hydrogeochemistry and sediment mineralogy of Lake Yanamate - anextremely acidic lake caused by discharge of acid mine drainage from the
Pb-Zn-(Cu) deposit, Cerro de Pasco (Peru), Securing the Future,
pp.1013-1022.

336

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

The Comparison of Monotonic Behaviors of Two Different Calcareous Sands
R. Rezvani, M.Sc. Student, School of Civil Engineering, Iran University of Science and Technology
H. Shahnazari, Assistant Professor, School of Civil Engineering, Iran University of Science and Technology
H. Salehzadeh, Assistant Professor, School of Civil Engineering, Iran University of Science and Technology
Y. Dehnavi, Ph.D. Candidate, School of Civil Engineering, Iran University of Science and Technology


ABSTRACT: Investigating the behavior of calcareous soils has been
extended during recent decades and the differences between themechanical behaviors of various calcareous soils due to their local
geological effects were mentioned by previous researchers. This
paper is containing an experimental investigation on the monotonic
behaviors of two originally different calcareous sands. These sands
were obtained from northern coast of Persian Gulf, including
Hormuz Island (HI) and Bushehr Port (BP). The testing program wasconsisted of two main parts: a series of isotropic compression (IC)tests to high stress levels (up to 1.5 MPa), and a series of drained
isotropically consolidated (CID) tests on reconstituted specimens.
Grain size analyses were carried out prior to and following of thetests to quantify the amount of particle breakage of used sands. The
effects of drastic parameters such as density and confining pressure
on compressibility and shearing response of the soils were also
investigated. The results showed that particle breakage occurred
during shearing and this phenomenon affected the mobilized friction
angle.

1. INTRODUCTION
Carbonate deposits which are located in tropical andsubtropical areas, cover approximately 40% of the ocean
floor [1]. Investigating on the mechanical behaviors of the
carbonate soils has been extended in a rapid rate in recent
years due to the existence of these deposits at many
petrochemical reserves and hydrocarbon industries such as
Persian Gulf. These reasons have encouraged many
researches to investigate the geotechnical behaviors of
calcareous sands [2]–[6]. Previous researches have showedthat there are many differences between the behaviors of
calcareous and terrigenous sands [7]–[9]. The most importantof these differences are related to the compressibility and thepotential of particle breakage of calcareous sands underapplicable pressures. Not only carbonate sands behave
different in comparison of terrigenous sands, but also various
types of calcareous sands have significant different behavior
in their response to the shearing loading due to their locations
and origins [9].
Particle breakage is one of the important features ofcalcareous sands and has a chief role on the mechanicalbehaviors of these sediments. For example, particle breakage
reduces the dilation response of the sands and mobilized
friction angle [9]. Many researchers try to quantify theamount of particle breakage of soils [10]–[12]. In somemethods, increasing of passing percentage of only one size ingradation curves of soils is analyzed which in someconditions may not be appropriate. Hardin (1985) used

overall gradation curves of soils to measure the particle
breakage with a good accuracy [13].
In this study, mechanical behaviors of two originally different
carbonate soils were investigated. The soils are obtained from
northern coast of Persian Gulf. Experimental research wasconsisted of two series of triaxial tests including a series ofisotropic compression (IC) tests and a series of drained
isotropically consolidated (CID) tests on reconstituted
specimens. In the isotropic compression tests, confiningpressure was increased up to 1.5 MPa to investigate thecompressibility of the sands under high pressure. To measure
the amount of particle breakage of the soils, the grain size
distribution analyses were carried out before and after thetests. Samples were prepared in various densities (i.e., loose,medium and dense) and consolidated under different
confining pressures to investigate the effects of theseimportant parameters on the shearing response of used sands.
2. SOIL CHARACTERIZATIONThe carbonate sands were obtained from the north of Persian
Gulf, Hormoz Island (HI) and Bushehr Port (BP). Hormuz
Island is located in the north of the famous Hormuz Strait.
Another soil was obtained from Bushehr Port which is one of
the most important and strategic commercial harbours of Iran.
Fig. 1 shows the origins of the used carbonate sands in thisstudy.
Grain size distribution tests showed that both carbonate sands
are well graded (SW). It can also be inferred that HI sand iscoarser than BP sand (Fig. 2).

Fig.1. Location of Hormuz Island and Bushehr Port in Persian
Gulf

337

Fig.4. FESEM micrograph for the BP sand;a: Subangular particles of soil; b: Close-up view of theparticles

Fig.2. Grain size distribution of the used sands
Also, Carbonate content tests using British Standard 1377,
were conducted on the used sands. The results showed that HI
and BP sands contain more than 53 and 44 percentages ofcarbonate calcium, respectively. Table.1 shows somephysical properties of the carbonate sands.
Table.1. Physical properties of the carbonate sands
Soil Type
Physical Property
HI
BP
AngularSubangularGrain Shape
0.78
0.43
D50 (mm)
8.33
7.87
Cu
1.54
0.84
Cc
14.48
13.21
γmin (kN/m3)
3
17.01
15.69
γmax (kN/m )
2.764
2.709
Gs
53.78
44.58
Carbonate Content (%)
Figs. 3 and 4 show the FESEM micrographs for HI and BP
soils. Microscopic photographs present that both types ofsoils are dominated by thin-walled mollusk and echinoderm
plate fragments and thick-walled foraminifera.

3. SAMPLE PREPARATIONIn this study, the specimens were prepared by the
air-pluviation method. The dry sand was spread into the moldwhich was fixed on the base pedestal of the triaxial apparatus
and lined with a rubber membrane. To acquire the desireddensity the mold was then tapped. The diameter and theheight of the specimen were 70 mm and 140 mm, respectively.
Saturation procedure contained flushing CO2, exuding
de-aired water and applying a back pressure of 200 kPa underan effective pressure of 10 kPa to the specimens. Following
this procedure, all specimens achieved high saturation
conditions, and the B-values greater than 0.95 insured thehigh quality of all the specimens. Afterward, samples were
isotropically consolidated under desired pressures. The
loading part of tests was conducted at strain control condition
with the rate of 0.5 mm/min in drained condition.
The tests were performed by the ELE International
Corporation Triaxial apparatus, located in GeotechnicalEngineering Research Center of Iran University of Scienceand Technology (GERC-IUST). Data Acquisition system
contained a PC machine which converting input signals to
bits through a DASH-12 analogue to digital converter (ADC).
The procedure of testing was managed and recorded with ageotechnical software called DataSystem7 (DS7). Table.2
shows the list of tests carried out in this study.
To measure the amount of particle breakage by the Hardin
method, grain size distribution test was carried out. Thismethod supposed that particle breakage ceases when thegradation curve of the soil reaches to the stable condition.
This condition was achieved when all particles become
smaller than 0.074 mm (sieve No. 200). According to theHardin method, grain crushing of the soil is defined as theratio of variation of grading curve to its all possible
variations:Br 
Fig.3. FESEM micrograph for the HI sand;a: Angular particles of soil; b: Close-up view of the particles

338

Bt
Bp

(1)

In which, Br is the relative breakage of the soil, Bt (total
breakage) is the area between the initial and the final grading
curves and Bp (breakage potential) is the area between theoriginal grain size distribution curve and the vertical line for
U.S. sieve No. 200 (Fig.5).
Test
Type
IC*
IC
IC
IC
IC
IC
IC
IC
IC
IC
CID**
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID
CID

Soil
Type
HI
HI
HI
HI
HI
BP
BP
BP
BP
BP
HI
HI
HI
HI
HI
HI
HI
HI
HI
HI
HI
HI
BP
BP
BP
BP
BP
BP
BP
BP
BP
BP
BP
BP

Table.2. List of tests
Confining
Relative Density
Pressure (kPa)
Condition
100
Loose
400
Loose
600
Loose
1000
Loose
1500
Loose
100
Loose
400
Loose
600
Loose
1000
Loose
1500
Loose
100
Loose
200
Loose
400
Loose
600
Loose
100
Medium200
Medium400
Medium600
Medium
100
Dense
200
Dense
400
Dense
600
Dense
100
Loose
200
Loose
400
Loose
600
Loose
100
Medium200
Medium400
Medium600
Medium
100
Dense
200
Dense
400
Dense
600
Dense

Fig.5. Definition of particle breakage based on Hardin
method (1985) [13]4. ISOTROPIC COMPRESSION TEST
Isotropic compression (IC) tests were performed on thespecimens of HI and BP calcareous sands which wereprepared in loose conditions. The confining pressure wasincreased up to 1.5 MPa and the volume change wasmeasured.
Fig. 6 shows the variation of void ratio of the two carbonate
sands under isotropic compression. It can be seen that the BP
sand has more compressibility than the HI sand. Particle
breakages of the samples were measured in isotropic
compression tests under various confining pressures. The
results showed that the BP sand has a higher particle breakage
potential than the HI sand. Particle breakage of the BP
samples is insignificant until the pressure of 600 kPa.
Afterward, this parameter increases intensely. For the HI sand,the main particle breakage occurs beyond 1000 kPa (Fig.7).
Hyodo et al (1999) and Kwag et al (1999) reported thisparameter for Aio, Dogs Bay, Amami and Quiou sand about
3000, 1000, 4000 and 900 kPa, respectively in medium anddense conditions [14], [15]. So, it can be concluded that, theparticle strength of the HI sand is higher than that of the BP
sand.

* Isotropic Compression** Drained Isotropically Consolidated Triaxial

Fig.6. Isotropic compression behavior of carbonate sands inloose conditions

339

showed no dilation and the volume decreases during thedeviatoric loading.

Fig.7. Particle breakage of samples in isotropic compression
tests in loose conditions
5. DRAINED ISOTROPICALLY CONSOLIDATEDTRIAXIAL TEST

Fig.8. Effect of confining pressure on the shearing strength of
the HI sand in loose conditions

A series of monotonic drained isotropically consolidated
triaxial (CID) tests were carried out to investigate theshearing behaviors of two originally different carbonate sands.
Several specimens in different relative densities (i.e., loose,medium and dense) were tested under various confiningpressures including 100, 200, 400 and 600 kPa. The
strain-controlled loading was applied at the rate of 0.5
mm/min and terminated when the axial strain reached to 20%.
In the theory of soil mechanics, the response of the usedcalcareous sands to the drained shearing loading contains twostages. At the first stage, samples tend to contract and theoverall volume of samples reduces (i.e., contactive behavior).
Continuing the loading causes the dilation response of thespecimens (i.e., dilative behavior). In this stage the volume ofthe samples increases.
In this study, the results showed that the BP sand had moretendency to volume change than the HI sand. In addition, at
high confining pressures, there was no dilation phase in both
HI and BP sands and only contractive phase occurred. Figs 8
to 11 show the drained shearing response of the HI and BP
sands in loose condition under varied confining pressures
between 100 and 600 kPa. The deviatoric stress in this study
was calculated based on the following relation:q  1 3

Fig.9. Effect of confining pressure on the shearing strength of
the BP sand in loose conditions(2)

It can be inferred from the Figs 8 and 9 that the shearing
strength of the HI sands is higher than that of the BP sands.
Figs 10 and 11 illustrate the volume change of HI and BP
sands. As it was seen during the isotropic compression tests,
the BP samples showed more volume change than the HI sandin similar loading conditions. For the HI sand, specimens
which were consolidated under 100 and 200 kPa had bothcontractive and dilative phases but in higher confiningpressures only contraction was observed. For the BP sand,only the specimen under the confining pressure of 100 kPa
had both contractive and dilative responses and the others

340

Fig.10. Effect of confining pressure on the volume change ofthe HI sand in loose conditions

Fig.13. Variation of mobilized friction angle of the BP
samples in drained triaxial tests

Fig.11. Effect of confining pressure on the volume change of
the BP sand in loose condition
The Mohr-Coulomb failure analysis was used to find themobilized friction angle of the specimens in different relative
densities (i.e. loose, medium and dense). Figs 12 and 13 show
the variation of mobilized friction angle of the HI and BP
sands in drained triaxial tests. Obviously, increasing theconfining pressure causes the decrement of mobilized friction
angle which was observed in this study too. The maximum
change in mobilized friction angle of the HI sand which took
place in loose conditions was 9.460 but for the BP sand theamount of this parameter was 12.30 and occurred in dense
conditions. The higher amount of variation in mobilized
friction angle of the BP sand can be related to the higher
amount of particle breakage of the BP sand in comparison to
HI sand which resulted in strength reduction.

Fig.12. Variation of mobilized friction angle of the HI
samples in drained triaxial tests

6. CONCLUSION
Experimental investigations were conducted to study themonotonic shearing behaviors of two calcareous sands. The
two originally different calcareous sands were obtained fromthe north coast of Persian Gulf. Experimental investigation
contained a set of isotropic compression tests (CI) and a set
drained isotropically consolidated triaxial (CID) tests.
Following results from the tests can be concluded:1) Initial tests about the physical properties of the usedcarbonate sands showed that the HI sand is coarser
and has greater maximum and minimum density.
These differences resulted from the different originsof the HI and BP sands with different geologicalcharacteristics.
2) Isotropic compression tests showed higher
compressibility index for the BP sand. In these tests,
the void ratio of the BP samples decreased at ahigher rate than that of HI samples.
3) It can be inferred from the isotropic compression tests
that the yielding stress of the HI and BP sands areabout 1000 and 600 kPa, respectively. Beyond thesepressures the amount of particle breakage increases
intensely.
4) Different behaviors in shearing response of the usedcarbonate sands were observed. The HI samples
showed more dilation than the BP ones. At highconfining pressures only contractive phase could beseen in both sands.
5) Particle breakage of the used sands caused to havemore reduction in mobilized friction angle byincrement of the confining pressure. Comparison
between the HI and BP sands, demonstrated that theBP sand has a higher reduction in mobilized friction
angle that was resulted from higher particle
breakage.

341

7. REFERENCES[1]

Holmes, A. "Principles of physical geology", Sunbury-on- Thames,
Nelson, London, 1978, pp730.
[2] Kaggwa, W. S., Poulos, H. G., and Carter, J. P. "Response of carbonate
sediments under cyclic triaxial test conditions", Engineering for
Calcareous Sediments, Proc. Int. Conf. on Calcareous Sediments,
Jewell and Andrews, eds., Balkema, Rotterdam, Perth, Australia, 1988,
pp. 97-107.
[3] Airey, D.W., "Triaxial Testing of Naturally Cemented Carbonate Soil",
Journal of Geotechnical Engineering, 1993, 119(9): 1379-1398.
[4] Salehzadeh, H., “The behaviour of non-Cemented and artificiallycemented carbonate sand under monotonic and reversed cyclic shearing,
University of Manchester”, U.K., 2000, Ph.D.
[5] Hasanlourad, M., Salehzadeh, H., and Shahnazari, H., "Dilation andparticle breakage effects on the shear strength of calcareous sands
based on energy aspects", International Journal of Civil Engineering,
2008, 6(2): 108-119[6] Dehnavi, Y., Shahnazari, H., Salehzadeh, H., and Rezvani, R.,"Compressibility and Undrained Behavior of Hormuz Calcareous
Sand", Electronic Journal of Geotechnical Engineering, 2010, Vol. 15.
[7] Datta M., Gulhati, S. K. and Rao, G. V., "Crushing of calcareous sands
during shear", Proceedings 11th Annual Offshore Technology
Conference, Houston, 1979, pp. 1459-1460[8] Coop, M. "The mechanics of uncemented carbonate sands",
Geotechnique, 1990, 40(4): 607-626.
[9] Brandes, H. "Simple Shear Behavior of Calcareous and Quartz Sands",
Geotechnical and Geological Engineering, 2011, 29(1): 113-126.
[10] Leslie, D.D., "Shear Strength of Rockfill", Physical Properties
Engineering Study No. 526, South Pacific Division, Corps of Engineers
Laboratory, Sausalito, Calif., Oct., 1975, pp. 124.
[11] Marsal, R. J., "Discussion of Shear Strength", Proceedings of the 6thInternational Conference on Soil Mechanics and Foundation
Engineering, Montreal, 1965, Vol. 3, pp. 310-316[12] Morrison, M.J., Mcintyre, P.D., and Sauls, D.P., "Lab- oratory test
results for carbonate soils from offshore African", Proceedings of theInternational Conference on Calcareous Sediments, Perth, Western
Australia, 1988, Vol. 2, pp.777-784.
[13] Hardin, B. O., "Crushing of Soil Particles", Journal of GeotechnicalEngineering, 1985, 111(10): 1177-1192.
[14] Hyodo, M., Aramaki, N., Nakata, Y., Inoue, S., Hyde and A.F.L.
"Particle crushing and undrained shear behaviour of sand", Cupertino,
CA, ETATS-UNIS, International Society of Offshore and Polar
Engineers, 1999.
[15] Kwag, J., Ochiai, H. and Yasufuku, N., "Yielding stress characteristics
of carbonate sand in relation to individual particle fragmentation
strength", Engineering for calcareous sediments, Edited by KA
Al-Shafei. AA Balkema, Rotterdam, the Netherlands, 1999, 79–87.

342

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Soaking Effect on Strength and Performance of Fine Grained Soil Stabilizedwith Recycled Gypsum
Aly Ahmed1, Keizo Ugai2 and Taira Jinno3
Civil Engineering Department, Beni-Suef University, Beni-Suef, Egypt
1, 2,3Civil and Environmental Engineering Department, Gunma University, Japan1



ABSTRACT: The use of recycled gypsum as a stabilizer material inground improvement projects has been initiated in Japan recently
since the disposing of gypsum wastes in landfill sites has manyenvironmental problems. As the use of recycled gypsum in soil
stabilization applications has many environmental and economical
benefits, it has many challenges since gypsum is soluble material.
Therefore, it’s essential to investigate the behavior of stabilized soil
with recycled gypsum under soaking conditions. Recycled gypsum
was mixed with Furnace cement type B with different ratios andvarious percentages of this admixture content were used. Cylindricalstabilized soil specimens were prepared and subjected to different
numbers of soaking periods. The soaked specimens were tested forcompressive strength, volume change and soil deterioration toinvestigate the effect of soaking conditions on their strength,
durability and performance. Results showed that the increase of
admixture content has a positive effect on the strength increase anddurability improves. The early soaking days have a negative effecton strength and durability compared to the later soaking days. The
increase of admixture content is associated with the increase ofpercentage of soil soluble deterioration. The use of Furnace cement
as a solidification agent to prevent the solubility and to improve thedurability of stabilized soil with recycled gypsum is recommended.
The influence of soaking conditions on strength, durability andperformance for specimens stabilized with the investigated limits ofrecycled gypsum in this study is limited. Generally, stabilized soft
clay soil with recycled gypsum, which is treated with Furnace
cement, has an acceptable durability as well as a good performanceagainst the actions of soaking conditions.

Keywords: Recycled gypsum; Soaking; Durability; Strength
1. INTRODUCTION
Gypsum waste plasterboard is a serious problem in Japan
since 1.6 million tons of gypsum wastes are generated
annually. The disposing of gypsum wastes in ordinary landfill
sites is not permitted in Japan according to Japaneseenvironmental regulations [1-4]. It must send to controlled
landfill sites to avoid the negative effect on the environment
and then the cost of their disposing increases [2]. Therefore,
investigations to produce recycled gypsum from gypsum
waste plasterboard and then to use as a stabilizer material inground improvement projects have been conducted recently
in Japan [1-3; 5-7]. In fact the use of recycled gypsum as astabilizer material has many economical and environmental
advantages but it has many challenges since gypsum is asoluble material. The solubility of gypsum is considered oneof the most challenges for the application of recycled gypsum,
produced from gypsum wastes, in ground improvement

projects. The solubility of gypsum results a negative effect on
both strength and durability since the bonding between soil
and gypsum is destroyed when water is introduced. To avoid
such negative effects, the additives of solidification agent
such as cement is essential to prevent the solubility of gypsum
as well to improve both strength and durability of soil
stabilized with recycled gypsum. The effect of soaking
conditions on soil stabilized with recycled gypsum has notbeen reported in literature up to the knowledge of authors. So,
it is essential for geotechnical engineers to have knowledge
about the effect of soaking conditions on the performance ofsoil stabilized with recycled gypsum to facilitate the use ofrecycled gypsum as a stabilizer material in ground
improvement projects. The influence of weathering or
environmental conditions in terms of freezing-thawing andwetting-drying cycles on the performance of soil stabilizedwith recycled gypsum have been conducted [9-11]. The
influence of freezing-thawing cycles on the durability of silty
sand soil stabilized with recycled gypsum was investigated.
Different contents of recycled gypsum-cement admixture
were mixed with tested soil in dry state; subsequently the
optimal moisture and dry density for each sample were
determined. The 7 days cured sample was subjected todifferent numbers of freezing-thawing and wetting-drying
cycles and then tested for compressive strength and durability.
Results showed that the actions of freezing-thawing had asignificant effect on strength and durability decrease
especially in the early cycles compared to the actions ofwetting-drying cycles. Samples stabilized with only recycled
gypsum did not survive against the actions offreezing-thawing and wetting-drying cycles. The additives of
furnace cement improved the durability and strength. The
performance and durability of stabilized samples wereenhanced with the increase of both recycled gypsum andfurnace cement contents [9-10]. The performance of soft clay
soil stabilized with recycled gypsum against the actions ofwetting-drying cycles was studied. For that purpose soft clay
soil was mixed with different contents of Bassanite andcement and then cylindrical stabilized soil specimens weremolded. Stabilized soil specimens were subjected to different
numbers of wetting-drying cycles to investigate their effecton strength and durability. Results showed that the early
cycles had a negative effect on durability and strengthcompared to the later cycles. Bassanite and cement contents
had a significant effect on the improvement of strength anddurability. As mentioned above, the main negative point for343

using recycled gypsum in ground improvement projects is thesolubility of gypsum when water is introduced. So, theinfluence of soaking conditions on soil stabilized withrecycled gypsum is considered the appropriate test in such
case.
This paper investigates the influence of soaking conditions on
the strength and durability of very soft clay soil stabilizedwith recycled gypsum and furnace slag cement type B. For
that purpose, recycled gypsum was mixed with furnace slag
cement type B with different ratios and then different
percentages of this admixture were added to tested soil. The
effect of admixture content and ratio on the performance andstrength of tested soil subjected to different soaking times isinvestigated.
2. MATERIALS AND METHODS
Three different materials are used in this research: very soft
clay, recycled gypsum and furnace cement type B. Soft clay
soil was brought from some construction site at Gunma
Prefecture in Japan. The initial water content for soil sample
was determined by an oven drying soil sample according to
ASTM specifications [12] and the average value was found160%. Properties for tested soil sample are presented in Table
1. According to unified soil classification, USC, system, thetested soil can be classified as clay soil with high plasticity(CH).
Table 1. Properties of tested soil
Property
Specific gravity, Gs
Water content, Wc %Liquid limit, LL %Plastic limit, PL %

Value
2.46
160
100
61.5

Property
Plasticity index
D60, mm
D50, mmD30, mm

Value
38.50
0.040
0.022
0.008

Recycled gypsum used in this research was produced fromgypsum waste plasterboard by heating process. Allprocedures for producing recycled gypsum from gypsum
waste plasterboard were presented in details in previous work[7]. Based on chemical analysis results for the produced
gypsum, the content of CaSO4.1/2H2O in the produced
recycled gypsum was found 92.6%. The recycled gypsum
was mixed with furnace cement at different ratios of 1:1, 2:1
and 3:1 for using this admixture as a stabilizing agent toimprove the performance and strength of tested soil. Fourdifferent percentages of these admixtures of 0, 7.5, 15 and22.5 % were mixed with tested soil to investigate theirinfluence on the performance of tested soil subjected tosoaking conditions.
Furnace slag cement type B used in this research was broughtfrom some cement company in Tokyo, Japan. This cement
type is mainly produced from waste materials and by-productof Portland cement manufacture. It has 30 to 60% of blast
furnace slag in its compositions in accordance JIS R5211
specifications [4]. It is important to evident that the main
purpose of using cement in this study is not only to improve
the strength of tested soil but also to prevent the solubility ofrecycled gypsum as mentioned before. Furnace cement wasadded to recycled gypsum with different ratios of 1:1, 2:1 and
3:1, respectively.
Cylindrical soil specimens with 50 mm in diameter and 100
mm in height were prepared. Firstly, recycled gypsum was
mixed in dry state with furnace cement at the investigateddifferent ratios of 1:1, 2:1 and 3:1, respectively. Subsequently,
soft clay soil sample was mixed with the desired
gypsum-cement admixture contents of 0, 7.5, 15 and 22.5%
according to testing program. The mixing process was doneby using an automatic mixer and mixing process wasprolonged for a certain time to ensure the gypsum-soil
mixture to be uniform. Secondly, the tested soil was placed inthree layers in special plastic mould, which is made for thispurpose. Each layer was compacted statically with efforts
equal to the efforts used in the standard compaction test. It is
important to evident that samples were compacted with their
original water content to simulate field conditions. Samples
were extracted from moulds after 24 hours. More attentions
were taken during sample preparation and extraction to
produce homogenous samples. For each test two to three
specimens were used and the average was considered. The
extracted samples were wrapped in polyethylene plastic sheet
and then kept in the controlled room at temperature 21±1 °C
for a certain time until required for the testing according totesting schedule. The molded samples were cured for 3, 7 and
28 days before to subject for soaking. For soaking test, thecured specimens were soaked in water for different intervaltimes of 0, 4, 7, 15 and 30 days and then tested for unconfinedcompression. The main target of soaking test is to investigate
the influence of soaking condition on the stability of soil
stabilized with recycled gypsum in terms of strength,
performance, durability and solubility. Besides, the effect of
curing time on the performance and strength of stabilized soil
specimens subjected to soaking condition is investigated.
Volume change and soil deterioration for tested soil samples
were reported after each soaking time.
3. RESULTS AND DISCUSSIONS
Aforementioned, the solubility of recycled gypsum isconsidered the main negative point for the use of recycled
gypsum in ground improvement projects. This property
results in deterioration for the developed bonding between
soil particles and gypsum and then the strength declines. The
use of any solidification material to prevent the solubility of
gypsum is essential needed to improve the performance andstrength of soil stabilized with recycled gypsum. Fig. 1 shows
the influence of curing time on strength ratio for samples
stabilized with 15% content of gypsum-cement admixture
and subjected to different soaking periods. Strength ratio isdefined as the ratio between the strength of sample subjectedto specified soaking time to the strength of identicalun-soaked sample. This figure indicates that the strength ratio
increases with the increase of soaking time in case of soaked
samples subjected to curing time of 3 and 7 days. The
increase of strength ratio in these cases is attributed to thefollowing reason. The chemical reaction between
gypsum-cement admixture and clay particles may need moretime to achieve the optimal strength, which might be morethan 7 days. Subsequently, the exposition of stabilized soil344

specimens in this case for soaking helped the specimens togain more additional curing. This additional curing time
promotes the reaction between gypsum-cement admixture
and clay particles to complete and then the strength ratio
increased with the increase of soaking time. After 30 days
soaking, the exposition of specimens which cured for 3 and 7
days for more soaking times has no significant effect on thestrength ratio increase. This is attributed to most of thereaction between clay particles and admixture is alreadycompleted within the 30 days of soaking. Thus there is no
more improvement in strength is developed. Besides, thesoaking of specimens for long time weakens the developed
cementations between soil particles. On the contrary forsamples cured for 28 days, the strength ratio decreases with
the increase of soaking time for specified time and after thatthe prolonged soaking time more than 15 days has
insignificant effect in strength ratio. The decrease of strength
ratio with the increase of soaking time in this case is attributed
the stabilized specimens after 28 days curing became almostin dry state. Subsequently, the exposition of dry specimens
for soaking destroyed the developed bonding force between
soil particles and then the strength decreased as presented
above.

soil structure and then the durability reduced. This result
agrees in concept with the results presented in the previous
works for studying the effect of wet-dry cycles and soaking
conditions on soil stabilized with different materials[10,13-15]. Secondly, the structure of stabilized specimens
subjected to soaking after a specified time, which is 15 daysin this case, may be able to rearrange itself again andaccommodates with the new environment. Subsequently, thedurability improved slightly after 15 days soaking due to theadditional gain of curing for stabilized specimens during
soaking time [10, 16]. This figure also indicates that, theadmixture content has a significant effect on the improvementof durability. This result is attributed to the increase of
gypsum-cement admixture content in soil mixture developed
enough hardening between soil particles which able toimprove the durability. The increase of admixture ratio isassociated with the decrease of durability as presented in Fig.
3. It is attributed to the increase of admixture ratio means thatthe decrease of cement proportion in admixture content. It is
well-known that the presence of cement in admixture content
is considered the main responsible factor to resist the actionsof soaking condition as well as improving the durability since
gypsum is soluble material.

3.50

1.20

Soaking after 3 days curing

3.00

Soaking after 7 days curingDurability Index

Strength ratio

1.00

Soaking after 28 days curing

2.50
2.00
1.50
1.00

0.80
0.60
0.40
Admixture content = 7.50%

0.50

0.20

Admixture content = 15.00%

0.00

Admixture content = 22.50%

0d

4d

7d

15d

30d

0.00
0d

Soaking time, days

The effect of admixture content and ratio on the durability ofstabilized soil specimens subjected to different soaking times
is presented in Figs. 2 and 3, respectively. Durability index isdefined as the ratio between the strength of sample cured for
28 days and then subjected to a specified soaking time to thestrength of identical un-soaked sample. The durability index
reduces with the increase of soaking time for all samples
stabilized with different contents and ratios ofgypsum-cement admixtures. It is clear that, the durability
index decreases significantly up to 15 days soaking andafterward the durability improves. This result is related to theeffect of early soaking times have a significant effect on
durability reduction compared to the later soaking times. This
result is attributed to the following reasons. Firstly, the rate ofwater absorption in the early soaking time is greater than that
occurred in the later soaking time. Subsequently, thereduction in durability increased with the increase of water
absorption rate. The penetration of water to stabilized soil
specimens results in disturbance for the developed
cementation between soil particles which results in change in

4d

7d

15d

30d

Soaking time, days

Fig. 2. Effect of admixture content on durability index forstabilized soil specimens subjected to different soaking times.
1.20
1.00

Durability Index

Fig. 1. Effect of curing time on strength ratio for stabilized
soil specimens subjected to different soaking times.

0.80
0.60
0.40
Admixture ratio = 1:1

0.20

Admixture ratio = 2:1
Admixture ratio = 3:1

0.00
0d

4d

7d

15d

30d

Soaking time, days

Fig. 3. Effect of admixture ratio on durability index forstabilized specimens subjected to different soaking times.
Deformation change in this study referred to settlement orswelling. It is the difference between the displacement345

measurements for the height of tested soil before and after tosubject for soaking. The positive value means swelling while
the negative value means settlement. The deformation change
was measured by using a micrometer through two fixed
points along the height of tested sample. The influence ofadmixture content at admixture ratio 1:1 on the deformation
change of stabilized specimens cured for 28 days and thensubjected to soaking condition is presented in Fig. 4. The
effect of soaking conditions on the swelling reduces after 15days soaking and then the samples start gradually to settle. It
is attributed to the presence of cement in the admixture
content resists the swelling phenomena because cement has ashrinkage property. Besides, the presence of cement insoil-gypsum mixture is solidified the activity of recycled
gypsum to swell. Subsequently, the chemical reaction
between recycled gypsum and clay particles, which causes
the swelling, is constrained and then the swelling is reducedwith the increase of soaking time. It is observed from Fig. 4
that, the effect of soaking conditions on volume change forsoil stabilized with recycled gypsum-cement admixture is not
significant since the maximum deformation change among all
tested samples was found less than 1.50 %. This proves thatthe use of recycled gypsum, which is solidified with furnace
slag cement, in ground improvement projects within theinvestigated limits in this study is durable against the effect ofsoaking actions in term of deformation change.
0.20

Vertical displacment, mm

0.15
0.10
0.05
0.00
-0.05

Admixture content = 7.5%

-0.10

Admixture conten = 15%
Admixture content = 22.5%-0.15(+) Swelling

The effect of admixture ratio on the deformation change ofstabilized specimens is presented in Fig. 5. As observed incase of admixture content, the swelling reduces gradually
after 15 days of soaking for all different investigatedadmixture ratios. This result is attributed to most of reactions
between gypsum-cement admixture and clay soil particles arealmost completed and afterward the potential of gypsum toswell is declined. It is obvious from this figure that, theincrease of admixture ratio increases the swelling. It is
attributed to the increase of admixture ratio means that theincrease of recycled gypsum proportion in soil mixture andthen the tendency of samples to swell is promoted.
The percentage of soluble soil deterioration particles forstabilized soil specimens with recycled gypsum wasdetermined based on the dry soil mass before and after
soaking. It is defined as the difference between the initial andfinal dry soil masses for soil sample before and after soaking
divided by the initial soil mass before soaking. Fig. 6 showsthe effect of admixture content on the percentage of soluble
soil deterioration for stabilized soil specimens subjected tosoaking conditions. The increase of soaking time increases
the percentage of soluble soil deterioration up to 15 days
soaking. Thereafter the percentage of deteriorated soil keeps
constant and there is no more soil deteriorated from
specimens. This result is related to most of water absorption is
occurred in the early soaking times. Water absorption isconsidered the major factor responsible about the increase of
the percentage of soil deterioration due to the change of soil
structure. This figure indicates that, the increase of admixture
content is associated with the increase of the percentage ofsoil deterioration. This result due to the increase of admixture
content in soil mixture increases the chance or possibility of
gypsum particles to dissolve in water since the amount of
gypsum in soil mixture is high and then the percentage of soil
deterioration is increased.

(-) Settlement

-0.20

6
4d

7d

15d

Soaking time, day

Fig.4. Effect of admixture content on deformation change forstabilized soil specimens subjected to soaking conditions.
0.20
0.15

Vertical displacment, mm

Admixture content = 7.5%
Admixture content = 15%
Admixture content = 22.5%30dSoulble soil particles, %

0d

0.10
0.05

5
4
3
2
1
0
0d

0.00
-0.05

Admixture ratio = 1:1-0.10

Admixture ratio = 2:1(+) Swelling

7d

15d

30d

Soaking time, day

Fig.6. Effect of admixture content on the deterioration ofsoluble soil from specimens subjected to soaking conditions.

Admixture ratio = 3:1
-0.15

4d

(-) Settlement

-0.20
0d

4d

7d

15d

30d

Soaking time, day

Fig.5. Effect of admixture ratio on deformation change forstabilized soil specimens subjected to soaking conditions.

Fig. 7 shows the effect of admixture ratio on the percentage ofsoluble soil deterioration for stabilized soil specimens
subjected to soaking conditions. The same behavior obtainedin case of admixture content is also obtained in the case of
admixture ratio. The increase of soaking time is associated346

with the increase of the percentage of soil deterioration fromtested soil samples. It is obvious from this figure that, there isno much difference between the effects of admixture ratios
investigated herein on the percentage of soil deterioration.
This result may be related to the suggested proportions of
cement in admixture content is enough to resist the actions ofsoaking condition. It is important to evidence that the tested
specimens did not subject for brushing process as in case ofwetting-drying test. This means that the target of the additionof cement in this case is only to prevent the solubility but it is
not to improve the strength. It is well-known that the increaseof cement proportion in admixture content is associated with
the increase of strength. While in case of soil deterioration,
the stabilized specimens did not subject to any external
stresses except their own weight. Thus, the minimum
suggested ratio of gypsum to cement which is 3:1 is capableto prevent the solubility of gypsum as well as to reduce thepercentage of soil deterioration.
6

Soulble soil particles, %Admixture ratio = 2:1
Admixture ratio = 3:1

4
3
2
1
0
0d

4d

5. ACKNOWLEDGMENT
Funding of this research is provided through a grant fromJapan Society for the Promotion of Science, JSPS to Dr.
Ahmed with ID No.: P10377. The authors appreciate thefinancial support from this organization.
6. REFERENCES

Admixture ratio = 1:1

5

for the enhancement of very soft clay soil is durable against
the actions of soaking in term of volume change.
5. Increasing soaking time increases the percentage of soluble
soil deterioration up to 15 days soaking and beyond thattime the percentage of the deteriorated soil is reduced.
6. There is no much difference between the effects ofadmixture ratios on the percentage of the deteriorated soil
for specimens subjected to different soaking times.
7. Furnace slag cement type B within the investigatedproportions in this study is recommended to use as asolidification agent for soft clay soil stabilized withrecycled gypsum.

7d

15d

30d

Soaking time, day

Fig.7. Effect of admixture ratio on the deterioration of soluble
soil from specimens subjected to soaking conditions.
4. CONCLUSIONS
Based on experimental results, the conclusion points can be
drawn as follows:
1. Both of admixture content and ratio have significant effecton the improvement of durability, strength andperformance of soil stabilized with recycled gypsum. The
durability index decreases with the increase of admixture
ratio.
2. The early soaking time has a negative effect on the
durability compared to the later soaking times and after 15days soaking the durability improved with the increase ofsoaking time.
3. The early curing times of 3 and 7 days have a positive effecton the improvement of durability and strength compared to
the later curing time of 28 days.
4. The effect of soaking conditions on the deformation change
for stabilized soil specimens is not significant since themaximum deformation percentage was found less than
1.50 %. This indicates that the investigated limits ofrecycled gypsum which is solidified with furnace cement[1] Ahmed, A., Ugai, K., and Kamei, T., “Laboratory and field evaluations ofrecycled gypsum as a stabilizer agent in embankment construction”
Journal of Soils and Foundations, vol. 51, December 2011, (In Press).
[2] Ahmed, A., Ugai, K., and Kamei, T., “Environmental evaluation forclayey soil stabilized with gypsum waste plasterboard in Japan” In theProceeding of the International GeoHunan 2011 Conference, June 9-11,
Hunan, China, GSP, ASCE, 2011, vol. 217, June 2011, pp. 9-17.
[3] Kamei T. and Shuku T., “Unconfined compressive strength ofcement-stabilized soils containing Bassanite produced from waste
plasterboard. Geotechnical Society Electronic J., vol. 2, March, 2007, pp.
237-244. (In Japanese)[4] Kamei T. and Horai H., "Development of solidification technology for
fluorine contaminated Bassanite using Portland blast-furnace (B)cement.” Japanese Geotechnical J., vol. 4, Jan. 2008, pp. 91-98. (InJapanese).
[5] Kamei, T., Kato, T., and Shuku T., “Effective use for bassanite as soil
improvement materials -Recycling of waste plasterboard-.”
Geotechnical Society Electronic J., vol. 2, March 2007, pp. 245-252 (InJapanese)[6] Ahmed, A., Ugai, K., and Kamei, T., "Application of gypsum waste
plasterboard and waste plastic trays to enhance the performance of sandysoil" In the Proceedings of Geo-Shanghai 2010 International Conference,
Geotechnical Special Publication, Ground Improvement andGeosynthetics, ASCE, vol. 207, June 2010, pp. 165-173.
[7] Ugai, K., Ahmed, A., “Evaluation of using gypsum waste plasterboard inground improvement.” In the Proceeding of the National Workshop on
Recycling Waste Plasterboard, Memorial Hall, Chuo University, Tokyo,
Japan, March 2009, pp. 9.
[9] Ahmed A., Ugai K. and Kamei T., " Durability of Fine Grained Soil
Stabilized with Gypsum Waste Plasterboard” In the Proceedings of 6thInternational Congress on Environmental Geotechincs 2010, New Delhi,
India, 8-12th November, vol.2, pp. 1469-1477.
[10] Ahmed A. and Ugai K., “Environmental effects on durability of soil
stabilized with recycled gypsum” Journal of Cold Regions Science andTechnology, vol. 66, May 2011, pp. 84-92.
[11] Kamei T., Ahmed A. and Ugai K., “Environmental effects on durabilityof soil stabilized with recycled gypsum” In the Proceedings of 2011
Pan-Am CGS Geotechnical Conference, October 2-6, 2011, Toronto,
Canada, p.10.
[12] Bowles, J.E., “Engineering properties of soils and their measurement.”
Text Hand book, 4th Edition, McGraw-Hill, New York, Inc. 1992, 241pp.
[13] Oti, J., Kinuthia, J., Bai J., “Engineering properties of unfired clay
masonry bricks.” Engineering Geology J., vol. 107, 2009, pp. 130–139[14] Masato, M., Hiroshi, T., Koji, K., “An experimental study on the
durability of fiber-cement-stabilized mud by repeated cycle test of drying
and wetting. Journal of the Mining and Materials Processing Institute ofJapan, vol. 121, February 2005, pp. 37-43 (In Japanese)347

[15] Khattab, S.A.A., Al-Juari, K.A.K., Al-Kiki, I.M.A., “Strength,
durability and hydraulic Properties of clayey soil stabilized with lime andindustrial waste lime. Journal of Al-Rafidain Engineering, vol. 16,
January 2008, pp. 102-116.
[16] Bin-Shafique, S., Rahman, K., Yaykiran, M., Azfar, I., “The long-term
performance of two fly ash stabilized fine-grained soil subbases”. J.
Resources, Conservation and Recycling, vol. 54, 2010, pp. 666-672.

348

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Heat Control of Pavement Surface Temperature Using Oyster Shell Lime
Satoru Ishiguro, Department of Bioresources, Mie University, Tsu, Japan
Masayoshi Yamanaka, Asahidoboku Co. Ltd., Yokkaichi, JapanABSTRACT: The developed heat reflective pavement is an
open-graded asphalt concrete in which voids in the upper part of thepavement are filled with a cement mortar containing oyster shelllime. In this study, the heat reflecting pavement using the oyster
shell lime was constructed in an open place, and the temperature
reduction of the pavement surface was evaluated experimentally bytaking measurements in summer. The temperature reduction effectsof the heat reflecting pavement were also investigated by laboratory
tests. The results show that the maximum surface temperature of thepavement fell by about 12°C compared with that of an asphalt
concrete pavement. It was found that the temperature reduction
effect was particularly due to the increasing reflectance of the solar
radiation on the pavement surface, which was indicated in the resultsof spectral reflectance tests.
Keywords: Heat reflecting pavement, Oyster shell lime,
Pavement surface temperature, Laboratory irradiation test,
Spectral reflectance
1 INTRODUCTION
Asphalt concrete pavements generally tend to absorb solar
radiation heat because they are black, and the temperature canrise to a maximum of 60°C or above at the surface in daylightin summer. The high temperature of the asphalt concrete
pavement will affect the durability of the pavement as well as
the heat island phenomenon in urban areas.
In Mie Prefecture, oyster farming is prosperous anddischarges a large quantity of oyster shell. Some of the shells
are pulverized in a factory after removing salt and utilized asoyster shell lime. Fig. 1 shows piles of oyster shell gathered ata factory site located in Toba city, Mie Prefecture. The
factory produces 6,000 tons of oyster shell lime per year.
In this study, a heat reflecting pavement using oyster shell
lime has been developed. The reduction of the pavement
surface temperature was evaluated by both laboratory tests
and field experiments.

2 MATERIALS AND PAVEMENT STRUCTURE
2.1 Pavement structure
A cross section of the pavement tested in this study is shownin Fig. 2. The thickness of the pavement is 50 mm. The heatreflecting pavement is an open-graded asphalt concrete inwhich voids in the upper part of the pavement are filled with acement mortar containing oyster shell lime. A maximum
aggregate size of 13 or 20 mm is used for the open-gradedasphalt concrete pavement. Fig. 3 shows the pavement
surface patterns. The beautiful patterns are produced bygrinding the pavement surface after the filling mortar has
hardened. The results of a skid resistance test and a wheel
tracking test show that the pavement has a good skid
resistance and plastic deformation resistance.
2.2 Materials
The oyster shell is a waste product from oyster farming. To
form the lime, the oyster shells are pulverized into a fineparticle size of less than 2 mm and used as the filling mortar
in this study. In the filling mortar mixture, Portland cement
and the pulverized oyster shells are used as a binder and a fineaggregate, respectively. As the oyster shell lime is lighter inFilling mortar with oyster shell lime

Voids

SubbasecourseFig. 2 Cross section of the pavement.

Fig. 1 Oyster shell piles gathered at a factory site
located in Toba city, Mie Prefecture.

Fig. 3 Pavement surface patterns.
353

Open-gradedasphaltcolor than the commonly used Portland cement, its inclusionin the mortar subsequently produces a mortar which is lighterin color. Consequently the surface of the open-graded asphalt
filled with the mortar increases the solar reflectance which
results in a lower surface temperature.

1.5

30.0

1.5

1.5
3.0
1.5

Table I Maximum temperature of the pavement surfaces

2009.8.11

Dense-gradedasphalt(°C )54.3

Heat reflecting
pavement
(°C )
45.0

Temperature
difference
(°C )

–9.3

2009.8.12

61.0

49.1

–11.9

2009.8.17

60.0

48.0

–12.0

2009.8.18

57.1

45.6

–11.5

Thermocouples (m)： Dense-gr aded asphal t： Unt reated： Water jet： Gr inding： Shot blastingFig. 4 Plan view of the pavement and surface treatments.
70

50
40
30

1000

Air temperature
Solarradiation

20
10
0

12 18
8/11

0

Temperature
difference(①－②)

6 12 18
8/12

12 18
8/17

0

6

Sola r radiation (W /m 2 )① Dense-graded asphalt② Heat reflecting pavement60

T e m pe ra tu r e （℃）

3.1 Field experiments
The pavements were constructed at the company site of the
branch office located in Ise city, Mie Prefecture in June 2009.
The temperature measurements were conducted in
mid-August 2009. Fig. 4 shows a plan view of the pavements
constructed at the outdoor testing site to monitor the surface
temperature in summer. The pavements are 50 mm thick andplaced on an asphalt binder course. An open-graded asphalt
concrete with maximum aggregate size of 20 mm was used
for the heat reflecting pavement. A dense-graded pavement
was constructed around the heat reflecting pavement as acontrol. After the filling mortar in the heat reflecting
pavement has hardened, the surfaces of the heat reflecting
pavement were subject to water jet, grinding, and shot
blasting treatments. The surface temperature was measured
by thermocouples attached to the pavement surfaces andrecorded by a data logger at 1 minute intervals. Temperaturesat depths of 25 and 50 mm from the surface were alsomeasured by thermocouples embedded at the same position in
the pavement. The air temperature and solar radiation weremonitored at the same time.
The temperature change of the heat reflecting pavement
surfaces are shown in Fig. 5. There was no rainfall during thisperiod. The black dense-graded asphalt rises to a maximum of
60°C at the surface, whereas it is not more than 50°C for theheat reflecting pavement. The surface temperature of the heatreflecting pavement on a fine day was up to 12°C lower thanthat of the dense-graded pavement. The temperature
reduction of pavement surface results from the higher solarreflectance of the surface filled with the mortar of a lighter
color. The maximum surface temperatures of the pavements
and the temperature differences are summarized in Table I.
Furthermore, results from tests conducted in August 2010
showed that the surface temperature difference between the
pavements was about 10°C. A decrease of the temperature
difference is partly because of the variation in the surface
colors of aged asphalt.

0
12 18
8/18

Date

Fig. 5 Temperature changes on the pavement surface.
70
60

T e m pe ra tu r e （℃）

3 EXPERIMENTS AND RESULTS① Dense-graded asphalt (25 m m)
③ Dense-graded asphalt (50 mm)② Heat reflecting pavement (25 mm)④ Heat reflecting pavement (50 mm)

50
40
30
Air temperature
20

Temperature
difference
(①－②)

10
0

12 18 08/11

Temperature
difference(③－④)

6 12 18
8/12

12 18
8/17

0

6 12 18
8/18

DateFig. 6 Temperature changes at depths of 25 and 50 mm.
354

3.2 Laboratory irradiation tests
Laboratory irradiation tests were also carried out to evaluate
the temperature reduction of the heat reflecting pavement
using the oyster shell lime. Specimens of five different
material types were used: a dense-graded asphalt, an
open-graded asphalt filled with a cement paste without oyster
shell lime, a (2:1) heat reflecting pavement, a (3:1) heatreflecting pavement, and an open-graded asphalt filled with amortar containing coral sand. The numbers in parentheses
represent the oyster shell lime and cement mix mass ratio for
the filling mortar. The (2:1) heat reflecting pavement is thesame as tested in the field experiments.
Fig. 9 shows a schematic of the arrangement for the lamp
irradiation tests. The specimens are 30 × 30 × 5 cm in size,
and the surface temperatures were measured by athermocouple embedded in the middle of the specimen. The
specimens were irradiated by a 150 W lamp for the first 3
hours and subsequently placed in a room at constanttemperature. The room temperature was approximately
27–28°C during the measurement.
Fig. 10 shows the surface temperature changes of thespecimens in the irradiation test. The difference at the highest
temperature between the dense-graded asphalt and the (2:1)
heat reflecting pavement was 12.4°C, and this value was thealmost same as the result of the field experiments exposed tonatural sunlight. The highest temperature of the open-gradedasphalt filled with cement paste was 47.6°C; 5.1°C higher
than that of the (2:1) heat reflecting pavement. The pavement
surface filled with cement paste is darker in color than thesurface filled with the mortar containing oyster shell lime.
The difference of the surface temperature is partially due tothe variation in surface colors.
Fig. 11 shows the surface temperature changes of thespecimens for different types of filling mortar. The highest
temperature of the specimen filled with mortar containing
coral sand is 43.9°C and is only 1.5°C higher than that of the(2:1) heat reflecting pavement. It seems that coral sand islighter in color and thus effective as a filling mortar. The
highest temperature of the (3:1) heat reflecting pavement is355

Fig. 7 Thermal image of the pavement.

Fig. 8 Visible image of the pavement.

Lamp

Thermocouple
5cm
5cm

Specimen(30 × 30× 5cm)
Expandedpolystyrene

Fig. 9 Arrangement for the lamp irradiation tests.
60
Dense-graded asphalt
Temp erature （℃）

The temperature changes at depths of 25 and 50 mm areshown in Fig. 6. The maximum measured daily internal
temperatures of the heat reflecting pavement fell about 10°C
compared with those of the dense-graded asphalt. The resultsof the field experiments show that the maximum temperatureof heat reflecting pavement decreases both on the surface andbelow the surface under summer sunlight irradiation, and thisacts to reduce the urban heat island effect.
Fig. 7 shows a thermal image of the pavement taken by an
infrared camera on August 17. Fig. 8 shows thecorresponding visible image. The temperature differences
between the heat reflecting pavement and the dense-gradedpavement are visually apparent in Fig. 7. The surface of theheat reflecting pavement filled with the mortar is significantly
lighter in color than the dense-graded pavement as shown invisible image (Fig. 8). Consequently, this leads to an increasein the solar reflectance and a lower surface temperature.

Open-graded asphalt
filled with cement paste

50

Heat reflecting pavement (2:1)

40

30
Room temperature
20
0

1

2

3

4

5
6
Tim e (h r)

7

8

9

10

Fig. 10 Surface temperature changes in the irradiation
test.

about 1°C lower than that of the (2:1) heat reflecting
pavement. The results show that a filling mortar with a highshell lime content achieves a slightly higher solar reflectance.
3.3 Spectral reflectance tests
A spectral reflectometer is used to evaluate the solarreflectance of the pavements. Specimens of four different
material types were used: a dense-graded asphalt, a (2:1) heatreflecting pavement, cement paste, and cement mortar
containing oyster shell lime. The specimens are 30 × 30 × 20

Temp eratu re （℃）

60
Open-graded asphalt filled withmortar containing coral sand50

Heat reflecting pavement (2:1)

40

mm in size.
Fig. 12 shows the spectral reflectance of the pavements.
The spectral reflectance in the near-infrared region is
0.05–0.11 and 0.55–0.7 for the dense-graded asphalt and heatreflecting pavement, respectively. The spectral reflectance ofheat reflecting pavement is significantly different from that ofthe dense-graded asphalt. In order to reduce the pavement
surface temperature, the results show that it is necessary toincrease the spectral reflectance, or solar reflection. Fig. 13
shows the spectral reflectance of the hardened cement mortar
and cement paste. The spectral reflectance in thenear-infrared region is 0.2–0.38 and 0.6–0.7 for the cement
paste and cement mortar, respectively. The values confirm
that the cement paste is darker in color and has a lower
reflectance compared with the cement mortar with oyster
shell lime.

Heat reflecting pavement (3:1)

4 CONCLUSIONS
30
Room temperature
20
0

1

2

3

4

5
6
Tim e (h r)

7

8

9

10

Fig. 11 Surface temperature changes of the specimens
with different types of filling mortar.

Reflectan ce (% )100Near-infrared region

80
60Heat reflecting pavement (2:1)

40

Dense-graded asph alt

20
0
400

700

1000

1300 1600 1900
Wav elength (n m)

2200

2500

Fig. 12 Spectral reflectance of the pavements.

5 ACKNOWLEDGMENT
The first author wishes to express his gratitude for thefinancial support of the Japan Society for the Promotion of
Science.

100
Near-infrared region

80
Reflectan ce (% )A heat reflecting pavement using oyster shell lime has beendeveloped and the temperature reduction effect was evaluatedby both laboratory tests and field experiments. The results ofthis study are summarized as follows.
1) The surface temperature of the heat reflecting pavement is
up to 12°C lower compared with the dense-graded pavement.
This is particularly due to the increase in the reflectance of
solar radiation on the pavement surface.
2) The measured spectral reflectance in the near-infrared
region was 0.05–0.11 and 0.55–0.7 for the dense-gradedasphalt and heat reflecting pavement, respectively.
3) The heat reflecting pavement and the open-graded asphalt
filled with mortar containing coral sand are lighter in colorand have a higher solar reflectance.
4) The beautiful patterns were produced by grinding thepavement surface after the filling mortar has hardened.
5) A large quantity of oyster shell is discharged from oyster
farming each year. Heat reflecting pavements using oyster
shell are environmentally friendly and contribute topromoting the effective use of a wasted resource.

6 REFERENCES
60[1]

Cement mortar containing oyster shell lime

40

[3]

20
0
400

[2]

Cement p aste

700

1000

1300 1600 1900
Wavelength (n m)

2200

2500

Fig. 13 Spectral reflectance of the cement mortar andcement paste.
356

The Society of Heating, Air-Conditioning and Sanitary Engineers ofJapan, Heat Island, Ohmsha Ltd., 2009, pp. 25-39.
Japan Society of Civil Engineers, Evaluation Technology of
Eco-friendly Pavement, Pavement Engineering Library 4, 2007, pp.
83-153.
Ishiguro, S. & Yamanaka, M. Research on Heat Reflection Pavement
Using Cement Mortar Containing Crushed Oyster Shell (in Japanese),
Proceedings of the Japan Concrete Institute, Vol. 32, No. 1, 2010, pp.
1367-1372.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Buckling Behavior of Composite Triangular Plates
T. Farsadi, Aerospace Engineering Department, Sharif University of Technology, Tehran, Iran
P. Amani, Aerospace Engineering Department, Sharif University of Technology, Tehran, Iran
E. Heydarnia, Aerospace Engineering Department, Amirkabir University of Technology, Tehran, Iran

ABSTRACT: This paper is to do a brief research on the buckling
behavior of composite triangular plates with various edge boundary
conditions and in-plane loads. It may be regarded as a right andsimple numerical method for the analysis of composite triangularthin plate using the natural area coordinates. Previous studies on thesolution of triangular plates with different boundary conditions were
mostly based on the Rayleigh-Ritz principle which is performed in
the cartesian coordinate. In this method, the energy functional of ageneral triangular plate is derived and the Rayleigh-Ritz method is
utilized to derive the governing eigenvalue equation for the bucklingproblem. The geometry is presented in a natural way by mapping aparent triangle and the integrals are evaluated analytically. The
polynomial terms Li , L j , Lk in the area coordinates are employed12

3

to interpolate plate deflection and La1 , Lb2 , Lc3 are used to enforce
the prescribe boundary conditions. Therefore, the admissible basic
functions given are L( i + a ) , L( j + b ) , L( k + c ) . In this approach, the1

2

Woinowsky-Krieger [7] derived the exact buckling load for a
simply supported equilateral triangular plate under anisotropic in-plane compressive load. Burchard [8] studied thebuckling of simply supported right-angled triangular plates
under uni-axial compression. There are several research
investigated the buckling of isosceles triangular plates undervarious loading conditions. Wakasugi [9,10] investigated thebuckling of simply supported and clamped equilateraltriangular plates. Conway and Lissa [11] searched thebuckling of equilateral triangular plates subjected to isotropic
in-plane compression. Tan et al. [12] and Tan [13] employed
the finite element method to study the buckling of generaltriangular plates. In the recent years, Wang and Liew [14]utilized their pb-2 Rayleigh-Ritz method to investigate thebuckling of isosceles and right-angled triangular plates underisotropic in-plane compressive load.

3

convergence is always assured due to the completeness ofinterpolating polynomials. Extensive buckling factors are presented
for several selected right-angled triangular plates of various edge
support conditions and subjected to composite thin plates undervarious in-plane compressive and shear loads and results arecompared with other available data..

Keywords: Triangular plate - Composite Material - Area
Coordinate - Buckling - Rayleigh-Ritz Principle

The basic aim of this research is to study and formulate theeffect of plate geometry aspect ratio and fiber orientation todetermine critical value of buckling load factor for acomposite triangular plate of various boundary conditionsand various in-plane (compression and shear) loads by using
an approximate-analytical solution for first critical mode
predominantly.
2 THEORETICAL ANALYSIS AND FORMULATION
2.1 Area-coordinate

1 INTRODUCTION AND BRIEF REVIEWIn many engineering applications, it is very important topredict buckling behavior of triangular plates. Buckling
phenomenon is critically dangerous to structural components
because the buckling of composite plates usually occurs at alower applied stress and generates large deformation. This led
to a focus on the study of buckling behavior in composite
materials. The use of composite materials in place of more
traditional isotropic materials has increased dramatically overthe past decades in areas such as the aerospace industry. With
the wide use of composite plate structures in modern
industries, dynamic and static analysis of plates of complex
geometry becomes an important part of engineering design.
Composite plates have been widely used due to their excellent
high strength-to-weight ratio, modulus-to-weight ratio andthe controllability of the structural properties with variationof fiber orientation. The problem regarding buckling of plates
under various shapes and boundary conditions has beenstudied in some famous classic and academic reference text
books [1,2] and some of numerous literature works [3-6].
It is interesting to note that buckling of triangular plates hasreceived far less attention than their rectangular counterparts.

In this section, the use of triangular coordinates to define theinterior of a triangular area is briefly described. This
description is important when formulating triangular finite
elements. The triangle in Fig. 1 is defined by coordinates(x i , yi ) of its vertices in the coordinate system O( x, y ) .
Cartesian coordinates are usually employed to describe any
point ( x 0 , y 0 ) on a rectangular area. They are, however,difficult to use when the boundary and the interior of atriangular area need to be defined. In order to define atriangular area, the coordinates ( x 0 , y 0 ) can be expressed in
the following way,x = x1L1 + x 2L2 + x 3L3
y = y1L1 + y2L2 + y 3L3where L 1 , L 2 , L 3 are triangular coordinates. The values of thetriangular coordinates range between 0 and 1 when thefollowing constraint is used,
L1 + L 2 + L 3 = 1(2)#Corresponding author (Touraj-Farsadi),
E-mail: touraj.farsadi@gmail.com, Tel: +905375213780

(1)

357

3( x 3 , y 3 )

y

y0

A1

A2
L2

3
1 3
ｶ
ｶ ｶLi
ｶ
bi
= 裹
=
2A i = 1 ｶLi
ｶx
i = 1 ｶLi ｶx
3
1 3
ｶ
ｶ ｶLi
ｶ
gi
= 裹
=
2A i = 1 ｶLi
ｶyi = 1 ｶLi ｶy
1 3 3
ｶ2
ｶ ｶ
ｶ ｶ
=
=
裹 b i b j ｶ L ｶL
ｶx ｶx
4A 2 i = 1 j = 1
ｶx 2
i
j
1 3 3
ｶ2
ｶ ｶ
ｶ ｶ
=
=
裹 g i g j ｶL ｶL
ｶy ｶy
4A 2 i = 1 j = 1
ｶy 2
i
j
1 3 3
ｶ2
ｶ ｶ
ｶ ｶ
=
=
裹 b i g j ｶ L ｶL
ｶx ｶy
ｶx ｶ y4A 2 i = 1 j = 1ijL1o( x 0, y 0 )2( x 2 , y2 )A3
L31( x 1, y1 )

O

x0

x

Fig. 1. Area coordinates in arbitrary triangular plate

Equation (1) together with (2) can be used to define any point( x 0 , y 0 ) in the triangle. The coordinates L 1 , L 2 , L 3 represent
dimensionless areas of triangles A1 , A 2 , A 3 as depicted in Fig.
1. These coordinates can be expressed as follows:

A1 + A2 + A3 = A(3)A
Li = i , i = 1,2,3
A
where A is the area of the triangle defined as,x1 x 2 x 3
2A = y1 y2 y 3
1

1

(4)

1

All expressions remain polynomial in the area coordinatesand can be easily integrated.
In other words,
鴿 ｶ2
・・ ｶ L2
・
1
・ ｶ2
・・ ｶ L2
鴿
鴿ｶ 2
・
2
・ｶ
・
・ ｶ2
・ｶ L
2
・
鴿ｶ
・
x
ｶ
・ 1
・ 2
・
・ ｶ L2
・ｶ
・ ｶ
・ｶ x
(8)
鴿
3
・
,
R
Q ・・
=
= 鴿
・
・
2
2
・
・
・ｶ L
・
・ｶ
2
y
ｶ
ｶ
・
・
・ 2
・
・
2
・ｶ L ｶ L ・
・ｶ
・ｶ y
・ 2ｶ
・
・
1
2
・
・
・ 2ｶ 2 ・
・ｶ L
x
y
ｶ
ｶ
・
・
3
・
・
・
・ｶ L ｶ L ・・ 2 3・
・ 2ｶ 2 ・・・・ｶ L ｶ L ・・ 1 3・
Where,

1 鴿・b1 b 2 b 3
鴿・R = 2A ・g g g
2
3
・・ 1
・ b2
b 22
b 32
・ 1
1
・ g2
鴿
g 22
g 32
・ 1
・Q =
4A 2 ・
・2b1g 1 2b 2 g 2 2b 3 g 3

When constraint (2) is accounted for, the triangularcoordinates can be solved from (5) as follows,Li =1(a + bi x + gi y ),2A i

i = 1,2,3

(7)

(5)

・

Where a i a coefficient that depends on is coordinates

(9)

(10)

・
b 1b 2
b2b 3
b 3b1
・
・
g 1g 2g2g 3
g 3 g1
・
・
b
g
g
b
b
g
g
b
b
g
g
b
+
+
+
( 1 2 1 2 ) ( 2 3 2 3 ) ( 3 1 3 1 )・

x j , yk of the three nodes in the coordinate systemO( x , y ) .
The coefficients can be expressed as follows:・

a i = x j yk - y j xk , bi = y j - yk ,g i = xk - x j , j , k = 1,2,3鴿鴿・R and ・Q are transfer matrices.

(6)

2.2 The displacement vector field
In the latter equation and below j, k and i form a cyclic
permutation.

For the displacement field in L1, L2 and L3 the area coordinate
system is used for the function as follows,Thin plate theory is used in the present analysis. However, toobtain the strains differentiation operator with respect toCartesian coordinates needs to be carried out. This is done asfollows,

w = w uw v
w u = w u ( L1, L2 , L3 ), w v = w v ( L1, L2 , L3 )(11)

w u ( L 1 , L 2 , L 3 ) is taken as the product of the boundary
equations and the internal line/curved support equations andw v ( L 1 , L 2 , L 3 ) is a polynomial in the area coordinates with358

an undetermined coefficient that the holder of thedisplacement field is expanded. The w u ( L 1 , L 2 , L 3 )function is defined,

w u = La1Lb2Lc3(12)

Where the value a, b, c depending on the free, simply
supported or clamped boundary conditions are 0, 1 or 2.
w v ( L 1 , L 2 , L 3 ) shape function includes terms with unknown
coefficients y ijk as follow:
p

w vI = 裹pp・ y ijk Li Lj Lk ,1 2

3

i = 0 j = 0 k= 0

(13)

i + j + k = p, 0 ｣ i , j , k ｣ p
So that the polynomial order p and i, j, k, respectively, display
the correct coordinates L1, L2, L3, varies from zero to p.
w v ( L 1 , L 2 , L 3 ) , shape function can be demonstrated withuncertainty factor that is just an index as follows. Such adisplay of computer programming is essential for the shapefunction, w v ( L 1 , L 2 , L 3 )N

w vI = ・ Y I L1L2L3
i

j

k

(14)

I=1

Where " I " is number of sentences and N is the count ofsentences. Performing some mathematical operations can beshown that a simple choice for the i, j, k using the following
equation,i = I - jN - kN 2 - 1
踐
I - 1 - kN 2 ・・
j = INT ・・・
・
(15)
・
・顏 N
踐
I -1
k = INT ・・ 2 ・・・・顏N ・second shape function is obtained by,combination of deflection functions, each of which satisfied
the rigid boundary conditions of the problem, and thenfinding the coefficients governing this linear combination byminimizing the total energy of the deformed body and thein-plane loads. Clearly, a proper choice of the deflection
expression is important to ensure good accuracy for the final
solution. The total energy is the sum of the strain energy ofthe plate, U, and the potential energy of the in-plane loads, V.
3 GOVERNING SYSTEM EQUATIONS OF PRESENTSYSTEM
In this paper, an elastic, composite, flat and thin triangularplate is used. The edges of the plate may be free, simply
supported or clamped supported. The boundary of this plate
defined by the equations L1 = 0, L 2 = 0, L 3 = 0 and theboundary conditions are provided by choosing a, b, c. It isnoteworthy that even if x and y axes are not coincided with themain purpose, there is no limit to the formulation. The
problem at hand is to determine the elastic buckling load ofsuch a general triangular plate subjected to an in-plane
compressive and shear load.
In accordance with established stiffness procedures, thevector of infinitesimal buckling strains consists of two parts.
(i) the linear flexural strain eL , and (ii) the nonlinear flexural
strain eN , that are given by,{ eL } = { eLx , eLy }・踐 2 踐 2・ ｶw ・ ｶw
・・ ,
= ・鉗
・鉗
・・鉗
顏ｶx ・ 鉐顏ｶy・{ eN } = { eNx , eNy , eNxy }

(16)

J=1

where,
e = J - fN - gN 2 - 1
踐
J - 1 - eN 2 ・・
f = INT ・・・
・
・
・
N
顏
(17)
踐
1
J
・
・
g = INT ・・ 2 ・・
顏N ・
and the Count sentences is possible by the following
equation:1
N = ( p + 1)( p + 2)
(18)
2
and thus specifying the order of the polynomial p and thenumber of sentence can get symbols i, j, k with number ofsentence" I ".
2.3 The Rayleigh-Ritz Method
The Rayleigh-Ritz method consists of assuming that somedesired deflection pattern can be approximated by a linear

T

(19)

T

・ ｶ2w ｶ2w 2ｶ2w
= ・・ 2 , 2 ,・・ ｶx ｶy ｶx ｶyNw vJ = ・ Y J Le1L2f Lg3

T

T

Infinitesimal buckling Linear and Nonlinear strains in an area
coordinate system are defined,{ e) L } = { eL , eL , eL }
1

2

T

3

・踐 2 踐 2 踐 2・ ｶw ・ ｶw ・ ｶw ・・・
鉗 ・・ ,
= ・・鉗
,・・
・・鉗
鉗
・ｶL3 ・
ｶL1 ・ 顏
ｶL2 顏
顏
・
)

T

{ eN } = { eN 1, eN 2, eN 3, eN 12, eN 23, eN 31 }・ ｶ 2w ｶ 2w ｶ 2w ｶ 2w
ｶ 2w
ｶ 2w
,
,
= ・
・ 2, 2, 2,・・ ｶL ｶL ｶL ｶL1ｶL2 ｶL2ｶL3 ｶL3ｶL1
1
2
3

Between vectors

(20)

T
T

{ eL } , { e)L } and { eN } , { e)N } using therelationship between Cartesian coordinates and the area
coordinates are expressed as the following:){ eL } = 鴿・R { eL }(21)){ eN } = 鴿・Q { eN }359

鴿So that transfer matrix 鴿・R , ・Q with fixed array is calculated・・

and available in (8).
3.1 Strain Energy
Adopting the classical thin plate theory, the strain energy U
due to bending is given in the matrix form, by,
T
1
U = ・
eN } 鴿
{・D { eN } dA
A2

(22)

Where "D" is the standard matrix of elastic constants
transformed to the local element coordinate system and isgiven by,鴿・d11 d12 0・鴿・D = ・d12 d 22 0・0
0 d 66・・

By substituting (29) and (30) into (27), strain energy inarea-coordinate is given by,

U = { Y}E1
t 3 v12E 2,d12 =
12 (1 - v12v21 )12 (1 - v12v21 )
3
E2
t
t3
d22 =
G
,d66 =
12 (1 - v12v21 )
12 12

U =

K IJ = ・

U = ・

) )
1 ) T 鴿eN } ・D { eN } dA
{
A2

)
T
鴿
鴿 鴿 鴿
・D = ・Q ・D ・Q(28)function w using the following (11), (12), (13) and (14) areavailable,

w I = w u w vI = < F I ( L1, L2 , L3 ) > { Y I } ,F I = La1+ i Lb2+ j Lc3+ k{ P }J(29)

that the arrays

(34)

of

matrix{P }if

3.2 Potential energy
The potential energy V of the orthotropic in-plane
compressive and shear load can be expressed in the matrix
form, as,

V = ・

T
1
eL } 鴿
{・N { eL }dA
A 2

(35)

While 鴿・N contains the in-plane compressive and shear loads

・in the manner,

鴿
・N x
鴿
N
=
・
・N
・・ xy

N xy
Ny

(36)

In the above expressions, Nx , Ny and Nxy are in-plane
compressive and shear load. By conversion load matrix intoarea-coordinate and substituting (21) and (36) into (35), thepotential energy V due to bending in area-coordinate is given
by,

V = ・

A

) )
1 ) T 鴿eL } ・N { eL } dA
{
2

(37)

where,

)
T
鴿
鴿 鴿
N = 鴿
・R ・N ・R
・

(30)

(33)

the exponent is characteristics a negative, to be chosen as
absolutely zero.

and for second shape function,

w J = w u w vJ = < F J ( L1, L2 , L3 ) > { Y J } ,F J = La1+ e Lb2+ f Lc3+ g){P }I 鴿・D { P } J dA
A

= { P (a ,b,c,e, f , g )} is obtained similarly.

( 6ｴ1)

Provided

where,T(32)

・ (a + i )(a + i - 1)L(a + i - 2)L(b+ j )L(c + k )・・
12
3・・(b + j )(b + j - 1)L(a + i )L(b+ j - 2)L(c + k )12
3・・・・(c + k )(c + k - 1)L(a + i )L(b+ j )L(c + k - 2)
{ P } I ( 6ｴ1) = ・・・ 2(a + i )(b + j )L(a + 1i )L(b+ 2j - 1)L(c3+ k- 1)・・1
2
3・・ 2(b + j )(c + k )L(a + i - 1)L(b+ j )L(c + k - 1)
1
2
3・・・・ 2(c + k )(a + i )L(a + i - 1)L(b+ j )L(c + k - 1)
1
2
3
・

(24)

(27)

(31)

where,t3

鴿 cos2 q
sin 2 q
- 2 cos q sin q
・
・
2
2
鴿
(25)
cos q
2 cos q sin q
・T = ・ sin q・2
2・cos q sin q - cos q sin q cos q - sin q
・and elastic constant matrix is obtained by,
T
鴿
鴿鴿鴿
(26)
・D = ・T ・D ・T
By transfer elasticity matrix into area-coordinate andsubstituting (21) and (26) into (22), the strain energy U due tobending in area coordinate is given by,)1
裹 K IJ Y I Y J
2 I J

So that,(23)

In the above expressions, E1, E2, v12, v21, and G12 are elastic
constants dependent on the physical characteristics of theplate material. For the composite material,

T・D { P } dA { Y }・ A 2 {P } 鴿In other words,

Where,

d11 =

1

T

(38)

function w using the following (11), (12), (13) and (14) areavailable,360

w I = w u w vI = < F I ( L1, L2 , L3 ) > { Y I } ,F I = La1+ i Lb2+ j Lc3+ k
and,w J = w u w vJ = < F J ( L1, L2 , L 3 ) > { Y J } ,F J = La1+ e Lb2+ f Lc3+ g(40)

By substituting (39) and (40) into (37), strain energy inarea-coordinate is given by,

V = { Y}TT )1・N { Z } dA { Y }・ A 2 {Z } 鴿range, 鴿・G is the geometric stiffness matrix corresponding to areference load, " l " is the eigenvalue, that is the load factor tomultiply to the reference load to obtain the critical value. As
is written, the equation system appears as an eigenvalues
determination problem, then for found some values of " l ",

(

)

鴿
det 鴿
・K - l ・G = 0(49)

4 SOME NUMERICAL RESULTS AND VALIDATION

1裹 G IJ Y I Y J
2 I J

So that,

G IJ = ・(48)

鴿・K is the stiffness matrix calculated in small-displacements(41)

In other words,

V =

鴿
(鴿
・K - l ・G ){ Y } = { 0}

(39)

T

(42)

)

{Z } I 鴿
・N { Z } J dA
A

(43)

where,

・(a + 1)L(a + i - 1)L(b+ j )L(c + k )・・
1
2
3・(b + j )L(a + i )L(b+ j - 1)L(c + k )(44)Z={ } I ( 3ｴ1) ・・
1
2
3・・(c + k )L(a + i )L(b+ j )L(c + k- 1)
12
3・・{ Z } = { Z (a ,b,c,e, f , g)} is obtained similarly.

Although the Rayleigh-Ritz method in area-coordinate
presented in this paper can be used to study buckling ofgeneral triangular plates, we will concentrate our analysis onright-angled orthotropic plates subjected to in-plane
compressive and shear load with various boundary conditions.
To describe the Boundary conditions in a composite thin plate,
we use letters F for free, S for simply supported and C forclamped edges. Consider an orthotropic triangular plate withlength a, a vertex angle , uniform thickness t, the modulus
of elasticity E1 , E2 , and Poisson's ratio v12 , v21 , as shown in
Fig. 2. This figure shows the selected right-angled composite
triangular thin plate studied in this paper.

J ( 6ｴ1)

B

It can be shown that when integrating polynomials in L1, L2and L3 over a triangle area A, the following relation holds,

a !b !c !
・ A L1L2L3dA = (a + b + c + 2) ! 2A
a b c

(45)

x2
y2

a0

b

A

x1
y1

x

3.3 Total Energy
Finally, with the function of both strain energy and thepotential energy of the total energy is written as,

P = U -V
(46)

(K I J Y I Y J - G I J Y I Y J )4.1 Validation
As the Rayleigh- Ritz method used in this study is anapproximate numerical approach, convergence andcomparison studies are essential to ensure the correctness andconvergence of the buckling results.

In other words,

1
裹
2 I J

y

Fig. 2. Composite right-angled triangular plateTT )1= { Y} ・P} 鴿{・D { P } dA { Y } A 2)
TT1{ Y } ・ A 2 {Z } 鴿・N { Z } dA { Y }

P =

a

x
C 3
y3

(47)

Table. I. Comparison of the buckling factor
N

3.4 Linear buckling analysis

N

Linear buckling technique allows one to obtain just thecritical load and the corresponding deformed shape of themodeled structure. To obtain this result, the condition ofneutral equilibrium between external loads and internal
reactions is searched, solving the equation,

361

b

(l

cr

=

b

a
N

Nb2 ) of the isotropic triangular plate
p 2D
2
1
2

a
Ref [14]
Present work345.82746.14252.63752.66698.69699.000

N

(l
N

b

a

2

cr

=

b

N

[10] Wakasugi. S., " Buckling of a clamped equilateral triangular plate underuniform compression", Trans. Japan. Soc. Mech. Eng. 1961, 27 (179),
1010-1017.
[11] Conway. H. D. and Leissa. A. W., "A method for investigating certain
eigenvalue problems of the buckling and vinration of plates", J. Appl. Mech.
1960, 27, 557-558.
[12] Tan, H. K. V., Bettess, P., Bettess, J. A. "Elastic buckling of isotropic
triangular flat plates by finite elements. Appl. Math. Modelling 7, 311-316(1983).
[13] Tan, H. K. V., "Elastic buckling of isotropic triangular flat plates byfinite elements", Proc. Inst. Civil Engineers 77, 13-21(1984).
[14] Wang , C. M., Liew, K. M., "Buckling of triangular plates under uniform
Compression" . Eng. Struct. 16, 43-50 (1994)

Table. II. Comparison of the buckling factor

Nb ) of the isotropic triangular plate
p 2D
2
1

a
Ref [14]
Present work

3

6.609
6.626

8.371
8.386

4.2 Results
Thebuckling

loads

are

expressed in terms of2
non-dimensional parameter, l = Nb , where N is theE 2t 3
critical buckling compressive or shear loads.
Figure.3 present the buckling factors for right-angled
triangular plates for different combinations of edge
conditions, fiber orientations and width-to-height ratios, b/a =0.5, 1, 1.5 and 2.

N

5 BRIEF CONCLUDING REMARKSNThis paper presents a research about the buckling ofcomposite triangular plates. The total energy functional for ageneral composite triangular plate is derived, and theeigenvalue equation for elastic buckling of the plate is
obtained through the application of the Rayleigh-Ritz method.
To this end, computer code has been developed to simulate
the buckling behavior of the composite triangular plate model.
Convergence and comparison studies are carried out to verify
the validity and accuracy of the solution method. Extensive
first-known buckling solutions for several selectedright-angled composite triangular plates are presented in thepaper. The influence of the fiber orientation on the buckling
behavior of the plates is examined. We find that the fiber
orientation and width-to-height ratios are most effective inbuckling behavior of the composite triangular plates.

baN(a) (CCC)(b(c) (SFS)(d

6 REFERENCES[1] Ugural A.C., "Stresses in Plates and Shells" , 2nd., WCB/McGraw-Hill,
1999.
[2] Timoshenko S., Woinowsky-Krieger S., "Theory of Plates and Shell",
2nd ed., McGraw-Hill, 1959.
[3] Tung T.K., Surdenas J., "Buckling of rectangular orthotropic plates under
biaxial loading", Journal of Composite Materials, Vol.21, No.2, 1987, pp.
124-128.
[4] Wang, J. T.S., Biggers S.B., Dickson J.N., "Buckling of composite plateswith a free edge in edgewise bending and compression", AIAA journal,
Vol.22, No.3, 1984, pp. 394-398.
[5] Farshad M., Ahmadi G., "Perturbation solution to the problems of generalorthotropic rectangular plates", Iranian Journal of Science and Technology,
Vol.1, No.2, 1971, pp. 147-162[6] Whitney J.M., Leissa A.W., Biggers S.B., Dickson, J.N., "Analysis of asimply supported laminated anisotropic rectangular plate" , AIAA journal,
Vol.8, No.1, 1970, pp. 28-33.
[7] Woinowsky-Krieger, S. "Berechnung der ringsum frei aufliegenden
gleichseitigen dreiecksplatte", Ing. Archiv, 1933. 4, 54-262.
[8] Burchard, W., "Beulspannungen der quadratichen platte mit schragseife
under druck bzw. Schub", Ing Archiv. 1937, 21(107), 332-348.
[9] Wakasugi. S., " Buckling of a simply supported equilateral triangularplate", Trans. Japan. Soc. Mech. Eng. 1960, 26, (164), 538-544.

362

3000

b/a = 2120b/a = 2

2700

100

2400
2100

80
1800

b/a = 0.5

b/a = 1.5
1500

60

b/a = 1.5

1200

40

900

b/a = 1600

b/a = 120
300

b/a = 0.5

0

0

10

20

30

40

50

60

70

80

0

90

 (deg)

0

10

20

30

40

50

60

70

80

90

80

90

 (deg)14018N

b/a = 2Nb/a = 1.5

120

16

b

N

14

a

b

N

100

a

12

80


b/a = 1640

4b/a = 120

20Nb/a = 1.560

8b/a = 0.5
010b/a = 0.5
20

30

40

50

60

70

80

0

90

 (deg)

0

10

20

30

40

50

 (deg)20
18
16

b/a = 2

14
12





N
10

10

b/a = 1.5

N(e) (FCF)

8
6

b

N

b/a = 1a4

b/a = 0.5

2
0

0

10

20

N

30

40

50

60

70

80

90

 (deg)Fig. 3. Variation of buckling factor
with respect to fiber orientation

for different width-to-height (b/a) and different boundary conditions
363

60

70

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Durability of Cement Treated Highway Materials
Sukolrat J., Klummeng A., Anuvechsirikiat S., Department of Highways, Thailand

ABSTRACT: Department of Highways, Thailand has used cement
modification to improve highway materials to overcome theproblems of scarcity of suitable materials. The modification
increases strength of local materials, which are not conformed thestandards for base courses and subbase courses in highway
structures. In some routes, cracks in road structures have been
observed on pavement constructed on cement modified structures
which may result from excess cement content. In order to prevent
such problem, durability of cement modified materials should be one
parameter to be concerned in addition to the strength to avoid excess
cement content in the design. The paper presents durability test
result of cement modified materials from six differrent locations
categorised as modified crushed rock and soil cement. The tests
include wetting and drying test, slake durability test, and wheel track
test to observe the performance in different durability testing. The
results show that cement content required for the modification canbe decreased to obtain satifactory durability, and materials’
suitability can be evaluated. A simple design chart for the cement
content selection can be used to pre defined the suitable materials for
cement modfication.

quantity of cement use, the quality control and quality
assurance during construction, and suitable design criteria. In
current design, the modified materials are subjected tostrength increment by cement content. Without consideration
for durability the materials may be unsuitable for roadstructure. This paper will present appropriated cement content
for modification by considering traffic capacity, durability of
cement modified materials, relationship between cement
content, strength of material and durability.

Keywords: Cement treared materials, highway materials,
durability, design chart

Fig. 1 Block cracks on asphaltic pavement
2. CURRENT SITUATION

1. INTRODUCTION
Land transportation is one important mode for both domestic
and international transportations. Thus, roadway is essential
for supporting economic and social development. With theincreasing usages of roads, new construction projects, roadwidening and maintenance are required. With suchcircumstance, road construction materials, conformed thequalification standards, are insufficient. Many standardisedmaterials need to be haulled from remote quarries costingover budget for the project. Besides, the haulage also affects
on service life reduction and pollutions. Modification ofengineering properties for local materials is one alternative
way to overcome such problems.
Most widely technique used for material modifiaction in
Thailand is cement modification. They have been used inroads construction for more than 30 years. Main purpose forsuch modifiaction is to strengthen materials for use as
basecourse and subbase course. Recently, the pavement
recycling technique has been used for reconstructed
deterirated pavement structure with cement mixing method.
However, reflective cracks occur on pavement surface insome routes after construction and the more damages
developed. Fig. 1 presents cracks on asphaltic pavement overcement modified crushed rock base.
The cause of cracks and failures of pavement on cement
modified materials structure may occur from inappropriate

Unsuitable aggregate is not allowed to use for road structure
to avoid undesired road performance. Currently, lack of road
construction materials has been underlined. So, it is necessary
to promote use of local materials for road structure withproperty improvement such as strength and durability toensure its performance for road standards.
Cement modified base course can increase ability for trafficload transferring, and decrease fatigue in sub base course. The
modification can also decrease possibility for cracks inpavement surface. Moreover, high strength increment in basecourse can decrease vertical deflection in sub base course andtensile strain in pavement surface on both dry and wet
conditions; hence, decrease opportunities in cracking andraveling. Besides, the base course can distribute traffic loadin larger area to underneath structures [1].
Normally, cement modified soil can be categorised into 5groups depending on cement content as follows [2]:2.1 Cement Modified Silty Clay
This type of modification was use in less cement which isapproximately 1-3 %. The objective is to modify silty clay
surface course, which contains high moisture and very soft,
to improve its strength and skid resistance.
2.2 Cement Modified Granular Soil
Cement modified granular soil require less cement content of
1-3 %. The cement can reduce plasticity and water absorption.
Although both cement modified silty clay and cement365

modified granular soil requires less cement content, themodification can improve properties clearly.

3. STUDY FOR DURABLE PROPERTIES FORHIGHWAY MATERIALS

2.3 Soil Cement
Soil cement has been designed for improve soils’
compressive strength and durability, so more cement content
is required. Normally, 5 % cement content is used for wellgrade soil. In the case that soil with high PI, lime is first added
to decrease its plasticity.

3.1 Highway Materials
Highway materials used in the research have been collectedfrom 6 different areas around Thailand. They can beclassified, according to highway’s standard as crushed rockand soil aggregate. Fig. 2 shows samples used in this study.

2.4 Plastic Soil Cement
Cement content in this kind of material should be enough forincreasing flow ability. The modified soil is always used forlining purpose i.e. ditch linings or side slope linings. Plastic
soil cement should be high workability. Normally it is also
used as grouting materials in concrete pavement repair.
2.5 Cement Treated Soil Slurry
Cement treated soil slurry contain cement, sand and water.
This material is the mixture of high water content with added
some admixtures for good workability. Normally it is used as
grouting material in concrete pavement repair similar toplastic soil cement.
Soil cement was firstly use in the United States inaccordance to cooperation research of South Carolina State
Highway Department and Portland Cement Association(PCA) to set up soil cement specification standard fordesigning and testing including moisture density test (ASTM
D558/AASHTO T134), wetting and drying test (ASTM
D559/AASHTO T135), and freeze and thaw test (ASTM
D560/AASHTO T136). All tests were used for samples
prepared with standard Proctor test to optimize cement
content at optimum moisture content and maximum dry
density. Unconfined Compressive Strength test (UCS) andsample preparation and curing followed ASTM D1632 andASTM D1633 [1].
Department of Highways, DOH, Thailand firstly applied
cement stabilization technique for road construction in 1965.
According to the standards, highway’s base course structuresuch as crused rock requires strength in term of CBR morethan 80%. However, in NortheasternThailand, there has beensign of lack for suitable materials for road construction.
Consequently, cement modified soil is one option formaterials used for highway’s structure.
The problem of soil cement modification in practice is micro
cracks [1]. Along with cement hydration, the structures arealways affected by volumetric shrinkage resulting inreflective cracks on road surface. The spacing between cracks
is approximately about 2.4-6 meters with crack width of lessthan 3 mm. Normally, cracks are very small and do not affectto traffic load capacity, and water cannot penetrate to lower
structure. However, excessive cement content will acceleratecracks causing damages for flexible road pavement.

a) Crushed Rock CR1b) Crushed Rock CR2c) Crushed Rock CR3d) Soil Aggregate AG1e) Soil Aggregate AG2f) Soil Aggregate AG3
Fig. 2 Samples used in the study
From compositions of each sample, CR1 is mainly composedof grey limestone. CR2 is basalt with high porosity. CR3 istaken from reclaimed base course which limestone is major
composition. For aggregates, AG1 comprises conglomerate
and breccias which contain quartz. AG2 is a mixture of sandand brown lateritic soil. AG3 mainly composes clay and silt
being weathered from mudstone.
Table 1 shows engineering properties of the samples
including gradation, liquid limits, plasticity index, optimum
moisture content, maximum dry density, CBR, swelling
index, and specific gravity. According to AASHTO
classification, CR1 can be classified as A-2-6(0) while CR2
and CR3 are both A-1-a(0). Aggregates AG1, AG2, and AG3
can be classified as A-2-6(2), A-2-4(0), and A-2-4(0)respectively.

366

Table 1 Engineerimg properties of samples
Sieve Analysis (%Passing)

Plasticity

Compaction

ID
3/4"

1/2"

3/8"

#4

#10

#20

#40

#100

#200

LL%

PI%

96.4
100
98.8
96.7
96.9
94.1

95.9
99.1
90.9
91.0
92.9
88.4

81.1
87.6
78.6
78.5
79.8
82.4

69.6
74.2
68.9
60.0
55.9
75.2

49.3
51.7
49.7
33.0
41.0
60.0

28.1
27.4
30.6
12.8
30.9
44.3

16.5
16.0
18.8
6.4
25.5
28.2

10.9
9.9
11.4
3.2
18.2
18.1

5.8
4.6
1.8
0.9
3.0
6.5

2.0
2.5
1.1
0.3
0.7
3.4

28.8
NP
NP
39.3
28.7
29.5

22.3
NP
NP
27.3
4.3
10.1

Strength of materials can be seen from CBR value. All
crushed rock samples show value of more than 80% while
aggregates show value of 20-30%. According to DOH’s
standard for crushed rock base course materials, CR1 has
high liquid limit and plasticity index. CR2 contains highproportion of fine material (finer than #200). CR3 has
gradation off specification. Aggregates from three locations
show that AG1 has high liquid limit while AG2 has high fineparticle content.
3.2 Strength of Cement Modified Materials
Samples are modified with cement. Crushed rock samples are
mixed with Portland cement Type 1 at 1%, 2%, 3%, and 4%
by weight, while soil aggregates are mixed with cement
content of 2%, 4%, 6%, and 8%. The samples are mixed at
OMC obtained from modified compaction tests. Samples areprepares in 4” proctor mold then cured for 7 days andsubjected for unconfined compression test. Fig.3 presents
results of unconfined compressive strength with respect tocement content for modified samples showing that higher
cement content results in higher strength as the effect of
cement hydration process.

dmax(g/cm3)
2.4
2.2
2.3
2.2
2.3
2.1

OMC
%
6
12
6
9
7.5
10

CBR%

Swell%

80
96
92
31.5
22.5
22

0
0
0
0
0
0

2.78
2.82
2.52
2.63
2.70
2.64

samples are broekn down into pieces about 40-60 g each. The
test follows ASTM 4644-08 for slake durability index. Fig.4
shows slake durability results for cement treated samples interm of slake durability index and cement content. The figure
shows that the higher the cement content, the higher
durability index as aggregates are stongly cemented.
Wetting amd drying test is one well-known method to
obtained durability. Samples were prepared under the same
cement content as strength test and were compacted using 4”
proctor mold and cured for 7 days. The test follows ASTM
D559-0. Fig.4 shows test result from wetting and drying test
for cement modified samples. Loss of weight is decreased as
the cement content incrases. From the figure, CR2 shows highloss of weight during wetting and drying cycle showing thatbaslt with high pososity has difficulty in cement modification.
100
95
90
Slake Durability Index, %

CR1
CR2
CR3
AG1
AG2
AG3

1"

GS

Lab CBR

70

85
80
75
70
AG1
65

AG2
AG3

Unconfined Compressive Strength (ksc)

60

CR1

60

CR2

55

CR3
50

50
0

1

2

3

4

5

6

7

8

Cement Content, %40

Fig.4 Slake durability test for cement modified materials

30
AG1
AG2

20

60

AG3

AG1

CR1

10

AG2

CR2

50

AG3

CR3

CR1

0
1

2

3

4

5

6

7

8

Loss of Weight, %0Cement Content, %Fig.3 Compressive strength of cement modified samples
3.3 Durability Test
Department’s standard is bound to strength of materials for
cement modified samples for pavement structure used.
However, durability is also a vital factor affecting
performance of cement treated materials. Two types ofdurability tests are carried out for each material: slake
durability and wetting-drying durability.
Slake durability test were performed on samples mixed at thesame cement content for strength test. After curing for 7 days,

367

CR2

40

CR3

30

20

10

0
0

1

2

3

4
5
Cement Content, %

6

7

8

Fig.5 Wetting and drying test for cement modified materials

4. PROPOSED DESIGN CRITERIA FOR
DURABILITY IN CEMENT MODIFICATION
Durability of cement modified materials is one factor
controlling performance of pavement structure. The
application for durability in cement modified materials has
been developed [3],[4]. From test results, samples can begrouped under durability criteria as Fig.6. The figure shows
relationship of loss of weight and unconfined compressive
strength.The loss is increased as the strength increased. With
the criteria from [3] and [4], the minimum cement content of
CR1, CR2, and CR3 are 2%, 3% and 1%, respectively. The
minimum cement content of AG1 and AG3 is 3% and AG2
requires at least 2% cement content.

The result shows slake durability index can indicate thesuitability of materials for cement modification. CR1, CR2,
and CR3 shows that slake durability index of modified
samples should be at least 70% and AG1, AG2, and AG3,
should have more than 55% slake durability index. Thismeans that the cement content should be high enough forincreasiin slake durability index to the recommended values.
5. PERFORMANCE TESTS FOR CEMENTMODIFIED MATERIALS
Performance of cement treated samples can be observed fromrepeatitive load test. Wheel track testing was used forobserving rutting depth of modified mateials. Samples wereprepared for CR1 at 0%, 1%, 2%, and 3% cement content, andAG1 at 0%, 2%, 4%, and 8%. Sample dimension is
305x305x100 mm and it was repetitively loaded under 20 kN
as shown in Fig.8.

Fig.6 Relationship between loss of weight and strength

Fig.8 Wheel track testing and rutting depth

Sample CR2 is clearly unsuitable for use as pavement
structure as showing high loss of weight. This material
requires high cement content for durability concept. In
addition, from the framework in Fig.6, for the pavement
structure design to sustain heavy traffic volume, strength ofcement modified materials can be possibly reduced to 14ksc(1.4MPa) for being in suitable durability range comparing to
the strength of 17.5ksc (1.75MPa) for current DOH’s design
standard.
Although wetting and drying test is a good indicator formaterial’s durability, one disadvantage is testing time. Hence,slake durability tends to be an alternative as it takes less time.
From test result, durability from both tests can be plotted as
Fig.7.

Both samples, CR1 and AG1, have high resistivity for rutting
when the cement contents increases as shown in Fig.9-10. At
10,000 passes, CR1 without cement modification shows
rutting depth of 6mm compared to 0.5mm with cement
content only 1%. In the same trend, rutting depth of AG1 isdecreased from 11mm to 0.5mm when 2% cement is added tothe samples. The test result implies that under repetitive loadcondition, small amount of cement used for modification isenough for resisting rutting on pavement structure.
6

Average Rut Depth, mm

5

4

3

CR1-Untreated2

CR1-1% Cement
CR1-2% Cement1CR1-3% Cement

0
0

2000

4000

6000

8000

10000

12000

Number of Passes

Fig.9 Rutting depth of sample CR1 against number of passes

Fig.7 Loss of weight and slake durabilty index of samples

368

12

Average Rut Depth, mm

10

8

6

4

AG1-Untreated
AG1-2% Cement
AG1-4% Cement2

AG1-6% Cement
0
0

2000

4000

6000

8000

10000

12000

Number of Passes

Fig.10 Rutting depth of sample AG1against number of passes

6. CONCLUSIONS
Cracks in cement modified structure have been reflected ontoasphaltic pavement when the cement content of suchmodified materials is not appropriate, quality of materials andquality control during construction are not suitable.
In accordance with durability concept, effect from unsuitablecement content can be underlined. The result from wetting
and drying test can be related to unconfined compressive
strength, normally used for design by the department. The
proposed design is to create guide for cement content required
for modification to obtain both enough strength and durability.
The study shows that, although CR2 can obtain high strengthfrom modification, it is no durable. So, the engineer can makejudment that CR2 may be inappropriate for use as cement
modified structure as high cement content will be required fordurability.
The result from wheel track test indicates that with small
cement content for modification, about 1-2%, can prevent
rutting on structure. Consequently, the engineer can possibly
decrease cement content to optimise compression strengthand durability properties because decreasing in cement
content can help to decrease possibility for reflective cracks
damages to the asphaltic structure.
Alternatively, in consideration of durability test, slake
durability test is quicker test than wetting and drying test. The
result shows the same trend on both methods. However, therelationship between results from both tests requires more
databases to obtain strong relationship for higher confidence.
7. REFERENCES
[1]
[2]
[3]

[4]

PCA, “Soil-Cement Information”, www.cement.com, 2003.
Ruenkrairerksa T, “Material modification and problems in Thailand”,
Department of Civil Engineering, KMITT, Bangkok, 1996.
Terrel, R. L., Barenberg, E.J., Mitchell, J.M. and Thomson, M.R., “Soil
Stabilization in Pavement Structures”: A User’s Manual, Vol.2:
Mixture Design Considerations, FHWA Report IP 80-2, Washington
D.C., 1979.
Ingles, O.G. and Metcalf, J.B., “Soil Stabilization Principle andPractice”, Butterworths Pty. Limited, Sydny-Melbourne-Brisbane,
Australia, 1972.

369

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Porous Geopolymer Concrete
Tawatchai Tho-in, Vanchai Sata, and Prinya Chindaprasirt
Sustainable Infrastructure Research and Development Center, Dept. of Civil Engineering
Khon Kaen University, Khon Kaen 40002, Thailand

ABSTRACT: This paper presents the use of fly ash based
geopolymer as cementitious materials in porous concrete. Porous
geopolymer concretes (PGC) were made of fly ash (FA), alkali
activators of sodium silicate (NS) and sodium hydroxide (NH)solution, and coarse aggregates. The PGC mix proportions used of
FA to coarse aggregate of 1:8 by weight, constant NS/NH ratio andliquid alkaline/FA (L/A) ratio at 0.50 and 0.45, respectively.
Crushed lime stone with 4.5-9.5, 9.5-12.5 mm diameter and three
concentrations of NH solutions at 10, 15, and 20 M were used. The
curing temperature of 60C for 48 h was used to activate thegeopolymerization. The results show that fly ash based geopolymer
binder could be used to made porous concrete with compressive
and splitting tensile strength range from 9.0-13.7 MPa and 1.1-1.8
MPa, respectively. The void content of PGC increased from 24.2 to
31.0% for water permeability coefficient from 1.09 to 2.58 cm/s.

Keywords: Porous concrete, geopolymer, sodium hydroxide,
fly ash, coarse aggregates
1. INTRODUCTION
Porous concrete or pervious concrete is a special concrete
which has relatively high water permeability as comparing
with conventional due to interconnected pore structure.
Typically porous concrete has void content between 18 to
35%, with compressive strengths of 2.8 to 28.0 MPa [1].
Porous concrete is traditionally used in parking areas, areas
with light traffic, pedestrian walkways, tennis courts,
greenhouse, and other civil engineering and architectural
applications [2-4]. In addition, it has environmental benefits
such as thermally insulating, acoustic absorption, and water
purification [5-6]. Porous concrete mixture is composed of
cementitious materials, coarse aggregate, little or no fineaggregates, water, possibly admixtures. Generally,
cementitious material used in porous concrete is Portland
cement.
Recently, the other form of cementitious materials in term ofgeopolymer or alkali-activated cementitious materials hasdeveloped. This cementitious material obtained from thegeoplymerization reaction of alumino-silicate oxide fromsource materials with high alkali solution [7]. The source
materials should be rich in silicon (Si) and aluminum (Al)from geological origin or by-product materials such as clays,
metakaolin, fly ash, bottom ash, and slag. The high alkali
solutions are usually sodium or potassium based solutions.
Many researchers [8-9] found that fly ash based geopolymer
can be developed as an alternative Portland cement withmany excellent properties such as high early strength,
excellent mechanical properties, and good resistance in acidand sulfate attacks. Furthermore, its use involves a lesser
amount of green house gas and environmental friendly
material compared to the conventional Portland cement [10].
This study focuses on application of fly ash based
geopolymer for making porous concrete. Lignite fly ash,sodium silicate and sodium hydroxide solution, and coarse
aggregate were used to produce porous geopolymer concrete(PGC). The compressive strength, splitting tensile strength,
void content, and water permeability of PGC were tested.

2. EXPERIMENTAL DETAILS
2.1 Materials
Lignite fly ash (FA) from Mae Moh Power plant in north ofThailand with mean particle size of 50 µm, Blaine finenessof 2250 cm2/g, and retained on sieve #325 at 45% wasselected as source materials. The particle shape of FA wasmainly spherical with the main chemical composition of
36.8% SiO2, 15.2% Al2O3, 19.7% Fe2O3, and 19.4% CaO.
Three concentrations viz., 10, 15, and 20 M of sodiumhydroxide solution (NH) and sodium silicate solution (NS)with 15.32% Na2O, 32.87% SiO2 and 51.81% water wereused as alkali activator. Crushed lime stone with 4.5-9.5 and9.5-12.5 mm diameter (assigned as CS and CM,
respectively) were used. The properties of all aggregates are
showed in Table 1.
Table 1. Properties of aggregates
Properties
Specific gravity
Water absorption (%)
Unit weight (kg/m3)
Abrasion resistance (%)
Fineness modulus

Aggregates (mm.)
4.5 – 9.5
9.5 – 12.5
(CS)
(CM)
2.72
2.71
0.21
0.21
1590
1570
31.1
33.2
6.06
7.00

2.2 Mix proportion and mixing
All PGC were made with fly ash to coarse aggregate 1:8 byweight. The NS/NH ratio and the liquid alkaline/FA (L/A)
ratio were kept constant at 0.50 and 0.45, respectively. The
typical mixture consisted of 221 kg/m3 FA, 33 kg/m3 NS, 66
kg/m3 NH, and 1768 kg/m3 coarse aggregate. All of PCG
mix proportions are shown in Table. 2. The symbol of
S-PGC15 concrete was PGC with CS aggregate and NH
concentration of 15 M.
The PGC were mixed in control room at 22-25 °C. First of all,
FA was mixed with NH for 5 minutes in a pan-type mixer.

371

Then, coarse aggregate was added and mixed for 4 more
minutes. Finally, NS was added and mixed for a last minute.
After mixing, fresh PGC were cast in cylindrical specimens
with 100 mm in diameter and 200 mm in height andcompacted using a vibrating table. The PGC specimens werewrapped with a thin plastic sheet and kept in the controlled
room at 25C for 1 hour. After that, they were placed in theoven for heat curing at 60oC for 48 hours. After the heatcuring, the specimens were put in the laboratory to cool down
and demoulded the next day. The specimens were wrappedwith vinyl sheet and stored in the controlled room at 23+2°C
and 50% RH until the testing age. Fig. 1 shows PGC
specimens after demoulding.
Table 2. PGC mix proportions
Materials (kg/m3)
Concrete
FA
NS
NH
CS
S-PGC10221
33.0 66.0 1768

following 24h exposure in air dry weight of sample (kg), W
is the density of water (kg/mm3), and V is the volume of thesample (mm3).
2.3.4 Water permeability
The water permeability of PGC was determined using theconstant head method following Darcy’s Law [14, 15]. Fig. 2
shows the schematic diagram of the experimental test set-up.
The coefficient of water permeability ( k ) is calculated asshown in equation (2).
QL
(2)
k
H At
Where: k is coefficient of water permeability (cm/s), Q isquantity of water collected (cm3) over time t (s), L is thelength of specimen (cm), H is the water head (cm), and A isthe cross sectional area of the specimen (cm2).

CM-

S-PGC15

221

33.0

66.0

1768

-

S-PGC20

221

33.0

66.0

1768

-

M-PGC10

221

33.0

66.0

-

1768

M-PGC15

221

33.0

66.0

-

1768

M-PGC20

221

33.0

66.0

-

1768

Fig. 2 Water permeability test set-up.

3. RESULTS AND DISCUSSION
Fig.1 Porous geopolymer concrete specimens
2.3 Testing details
2.3.1 Compressive and splitting tensile strength
The compressive strengths of PGC were determined at 7 daysin accordance with ASTM C39 [11]. Before testing, thecylindrical specimens were capped at both ends with a sulfur
capping compound to level the loading surface. The splitting
tensile strengths at 7 days were measured by performing inaccordance with ASTM C496 [12].
2.3.3 Void content
The total void ratio of PGC was determined by taking thedifference between the weight of the cylinder sample insaturated under water and air dry sample [13]. The equation
used to calculate this value is shown in equation (1).

3.1 Compressive and splitting tensile strength
The density, compressive strengths, and splitting tensile
strength of PGC are shown in Table 3. Because of the highporosity of PGC, the density of all PGC ranged from 1750 to1860 kg/m3 which were lower than conventional concrete by
about 25%. The compressive strengths of PGC were lower
than those of normal Portland cement concrete. However,the fly ash geopolymer could be applied to produce porous
concrete with acceptable compressive strength of 9.0 to 13.7
MPa. The splitting tensile strengths varied between 0.9 and
1.4 MPa.
Table 3. Compressive and splitting tensile strength of PGC
DensityCompressive
Splitting tensile
Concrete(kg/m3)strength (MPa)strength (MPa)S-PGC10
13.6
1.6
1860
S-PGC15

1800

13.7

1.8

S-PGC20

1810

11.9

1.5

M-PGC10

1840

11.4

1.2

Where: VC is the void content of PGC (%), W1 is the

M-PGC15

1800

12.0

1.4

saturated under water weight of the sample (kg), W2 is the

M-PGC20

1750

9.0

1.1

VC  [1  (W2  W1 ) / W V )]  100

(1)

372

Fig. 3 shows the relationship between NH concentration andthe compressive strengths of PGC with CS and CM
aggregates. At same NH concentration, the PGC containing
CS aggregate had compressive strength higher than those of
CM aggregate. For example, the compressive strength of
S-PCG15 was 13.7 MPa, while M-PGC15 was 12.0 MPa.
This is due to the PCG with small size of coarse aggregate
was due the larger specific surface area of CS particle ascomparing with the CM which helped in higher contract areaof the aggregate particles and paste, resulted in high bonding
and compressive strength [16].
The effect of NH concentrations between 10 and 20 M on
PGC compressive strengths showed that the optimum NaOH
concentration to produce good strength PGC was 15 M.
Because of the formation of gel depended on theconcentration of alkali OH ion, the low NH concentration
resulted in a weak chemical reaction, and the compressive
strength was low. An increase in alkali concentration
enhanced strength development of geopolymer, but excesshydroxide ion concentration caused aluminosilicate gel
precipitation at early stage, resulting in lower strength
geopolymer [17, 18]. A maximum compressive strength of
13.7 MPa were obtained at S-PGC15 concrete.

Fig. 3 Compressive strength of PCG
The effects of NH concentrations and aggregate sizes on
splitting tensile strength of PGC are shown in Fig. 4. The
results had the same direction of the compressive strength.
The PGC containing CS aggregate had the splitting tensile
strength higher than those of CM aggregate. The use of 15
M NH in PGC had the highest splitting tensile strength.
Similar to the compressive strength, S-PGC15 concrete gave
the maximum splitting tensile strength at 1.8 MPa. The ratioof splitting tensile to the compressive strength ranged from
10.6 to 13.0% with the average of 12.0%. These ratios wereslightly higher than 10% for the conventional concrete. Thiswas probably due to the stronger interfacial transition zone
between aggregate and gepolymer binder compared to thatof Portland cement binder [19].
3.2 Void content and water permeability
From this study, the void content of PGC varied between 24.2
and 31.0% with average value of 26.8% depended on coarse

aggregate size and NH concentration. The relatively highvoid contents of PGC led to high water permeability
coefficients which ranged from 1.09 to 2.56 cm/s. All of void
contents and water permeability coefficients of PGC aresummarized in Table 4.

Fig. 4 Splitting tensile strength of PGC
Table 4. Void content and water permeability of PGC
Water permeability
Void content
Concrete
coefficient ( k ) (cm/s)(%)
S-PGC10

24.2

1.16

S-PGC15

25.3

1.09

S-PGC20

27.4

2.00

M-PGC10

24.9

1.73

M-PGC15

28.0

1.79

M-PGC20

31.0

2.58

The influence of coarse aggregate sizes and NH
concentrations on void content of PGC are shown in Fig. 5.
It was showed that the PGC containing CS aggregate had
lower void content than those of CM aggregate for allconcentrations of NH. Since, the small size of CS aggregate
had the unit weight higher that of CM aggregate, it had
lower void between its particles. In addition, it was foundthat the M-PGC20 concrete with the highest void content of
31.0% had the lowest density of 1750 kg/m3, and,oppositely, the S-PGC10 concrete with the lowest void
content of 24.2% had the highest density of 1860 kg/m3.
The relationship between NH concentration and water
permeability coefficient of PCG with CS and CM aggregate
are shown in Fig. 6. The trend here was similar to the void
content results. At the same NH concentration, lower valuesof water permeability coefficient were achieved in PGC
using small aggregate size. Also, the water permeability
coefficient of PGC obtained from different NH
concentration. Thus, the relationship between void content
and water permeability of PGC could be fitted in exponential
curve as shown in Fig. 7.

373

3. The void content of PGC ranged from 24.2 to 31.0%.
This high porosity leaded to a high water permeability
coefficient at 1.09 to 2.58 cm/s.

5 ACKNOELEGMENT
This work was supported by the Higher Education Research
Promotion and National Research University Project ofThailand, Office of the Higher Education Commission,
through the Advanced Functional Materials Cluster of Khon
Kaen University.
Fig. 5 Void content of PGC

6 REFERENCES[1]

Fig. 6 Water permeability of PGC

Fig. 7 Relationship between void content and water
permeability coefficient of PGC

4 CONCLUSION
From the test results, the following conclusions can be drawn
as follow:
1. Fly ash based geopolymer could be used as
cementitious materials for making porous concrete withacceptable compressive strength of 9.0 to 13.7 MPa.
2. The compressive and splitting tensile strengths were
optimum with PGC containing small aggregate size
and NH concentration of 15 M.

ACI committee 522, Pervious Concrete, Report No. 522R-10,
American Concrete Institute, Detroit, USA, 2010 P.38.
[2] Schaefer VR, Wang K, Sulieman MT, Kevern JT. ”Mix design
development for pervious concrete in cold weather climates.” Final
Report, Iowa Department of Transportation, National Concrete
Pavement Technology Center, Iowa Concrete Paving Association,
February 2006, P.67.
[3] Scholz M, Graboweiki P. “Review of permeable pavement system,”
Bulid. Environ., 2007;42:3830-36.
[4] Tyner JS, Wright WC, Cobbs PA. “Increasing exfiltration frompervious concrete and temperature monitoring,” J. Environ. Manage.,
2009;90:2636-41.
[5] Park AB, Tia M. “An experimental study on the water-purification
properties of porous concrete,” Cem. Concr. Res., 2004;34:177-184.
[6] Kim HK, Lee HK. “Acoustic absorption modeling of porous concreteconsidering the gradation and shape of aggregate and void ratio,” J.
Sound Vib., 2010;329(7):866-79.
[7] Davidovits J. “Geopolymer: inorganic polymeric new materials,” J
Therm Anal 1991;37:1633-56.
[8] Hardijito D, Wallah SE, Sumajouw DMJ, Rangan BV. “On thedevelopment of fly ash based geopolymer concrete,” ACI Mater J.,
2004;101(6):467-72.
[9] Bakharev T. “Resistance of geopolymer materials to acid attack,”
Cem Concr Res., 2005;35:658-670.
[10] Duxson P, Provis JL, Lukey GC, van Deventer JSJ. “The role ofinorganic polymer technology in the development of ‘green concrete,”
Cem Concr Res., 2007;37:1590-7.
[11] American Society for Testing and Materials (ASTM) C 39/C 39M-01.
Standard test method for compressive strength of cylindrical concrete
specimens: 2003.
[12] American Society for Testing and Materials (ASTM) C 496-96.
Standard test method for splitting tensile strength of cylindricalconcrete specimens: 2003.
[13] Park AB, Tia M. “An experimental study on the water-purification
properties of porous concrete,” Cem Concr Res., 2004;34:177-184.
[14] Japan Concrete Institute, Proceedings of the JCI Symposium on
Design, Construction and Recent Applications of Porous concrete(2004) pp.1-10.
[15] Park SB, Lee BJ, Lee J, Jang YI. “A study on the seawater
purification characteristics of water-permeable concrete using
recycled aggregate,” Resour Conserv Recy., 2010;54:658-65.
[16] Al-Oraimi SK, Taha R, Hassan HF. “The effect of the mineralogy ofcoarse aggregate on mechanical properties on high-strength concrete,”
Constr Build Mater., 2006;20:499-503.
[17] Somna K, Jaturapitakkul C, Kajitvichyanukul P, and Chindaprasirt P.
“NaOH-activated ground fly ash geopolymer cured at ambient
temperature,” Fuel., 2011;90:2118-24.
[18] Lee WK van Deventer JSJ. “The effect of inorganic salt
contamination on the strength and durability of geopolymer,” Colloids
Surf A., 2002;211:115-26.
[19] Sarker PK. “Bond strength of reinforcing steel embedded in fly
ash-based geopolymer concrete,” Mater Struct., 2011;44(5):1021-30.

374

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Design and Fabrication of Composite Technology for Earth Slope Protection
Md. Toriqul Islam , Department of Sustainable Social Systems, Mie University
Md. Zakaria Hossain, Department of Environmental Sci. and Tech., Mie University
Masaaki Ishida, Department of Sustainable Social Systems, Mie University


ABSTRACT: This paper presents the design and fabrication ofcomposite technology for earth slope protection. The selectedstructures that have been identified for earths slope protection are
skeletal frame of soil anchoring, earth reinforcements, hill cover andslope surface lining. Design sketches of various composites alongwith fabrication procedure are depicted. Some examples of practicalapplications of this composite technology have been demonstratedusing simple techniques. Based on the cost of required materials andlabor, the cost of fabrication per unit area of composite structure has
been calculated. It is concluded that the required cost of thecomposite structure was U$ 60 dollar per square meter.

Keywords: Composite structure, Design, Fabrication, Slopeprotection, Application.
1. INTRODUCTION
Composite technology consists of sand-cement mortar
reinforced with wire mesh are used in the construction ofvarious structures [1]. It has also been used for making linings
to bridge arches, swimming pools, flumes, aqueducts, thinshell structures and low-cost farm structures [2]. Researchand development work on this material in Japan, however, arenot so many. Only a few research institutions in Japan like
Building Research Institute, Tokyo Institute of Technology,
Nihon University including the Mie University and Kyoto
University have applied the composite techniques but mostly
limited to demonstration only [3-7]. In view of the above
distinct advantages of this material, it may be effective toutilize this technology for earth slope protection in the hilly
country like Japan. Understanding the facts stated above, a
study was undertaken in the Division of Environmental
Science and Technology, Mie University, Japan with thefollowing objectives. i) to identify the structures that can bemade with composite technology for earth slope protection,and ii) to present the fabrication process and cost estimationof composite structures for earth slope protection.
2.

MATERIALS AND METHODSIn the field of earth slope protection, the basic materials
needed for composite constructions are wire mesh, sand,cement and water. The skeletal steel is used in theconstruction of composite frame structures for earth slope
protection. A basic description of the constituent materials
and construction procedure is given below.
2.1 Constituent Materials
Wire mesh: As given above, there were different types ofmeshes used in composite works. It is one of the mostessential components of composite especially for slope
surface protection works. The wire meshes must be easily
handled and flexible enough to be bent around sharp corners
and unevenness of the slope surface. The main function of thewire mesh in slope surface protection is to act as a lathproviding the form and to support the mortar during thehardened state of composite. The wire meshes absorb thetensile stresses on the structure which the mortar, on its own,
would not be able to withstand. In the project special types ofsteel wire mesh (galvanized metal mesh) having 2.0 mm wire
diameter with center to center wire spacing of 50 mm wereused. The mesh (Fig.1) has been folded for the ease oftransportation and can easily be spread during construction. It
has been coated with zinc to protect from rusting according to
the Japan Industrial Standard (JIS G3552).

Fig. 1 Mesh used for composite technology for slope
protection
Cement: The ordinary Portland cement is the most commonlyused in Japan among several types of cement available
commercially. In the present study, the ordinary Portland
cement (JIS R5210) was used throughout the research work.
This type of cement is adequate for application in earth slope
protection where special conditions do not prevail.
Sand: For composite work, well graded coarse sand iscommonly used. In the present study, the river coarse silica
sand is used. The sand is passing through 5.0 mm sieve andthe fineness modulus of it is 2.5.
Water: Water plays a vital role on the resulting hardened stateof composite. Sea water or water with any impurities such asacids, soluble salts and any organic matters are not at all
suitable for mixing mortar as they may increase the risk ofcorrosion or interfere the setting time of cement and finally375

the strength of the composite. In this study tap water was used
for making the cement mortar.
2.1 Construction Procedure
Placement of wire mesh: For placement of wire mesh, noskeletal steel frame is needed for the composite work in earth
slope protection. The slope is cleaned after removing the trees,
grass and any other element before placing the mesh. An
example of this procedure is shown in Fig.2.

2.2 Preparation of Mortar
In order to obtain the desire strength of the structures, it isvery important to select the suitable proportion of the cement
to sand ratio.For the composite in building works, theproportion of cement to sand generally varies from 1 part ofcement to 1.5 to 2 parts of sand by weight. However, for thecomposite in slope protection, the proportion of cement tosand is taken as 1 part of cement to 3 to 5 parts of sand by
weight. This provides the strength of more than 15 N/mm2
which is considered enough for slope protection. The water tocement ratio used was in the order of 0.45 depending upon thedryness of sand. In the mixing process the sand and cement
were firstly mixed uniformly. The water was added graduallypart by part in order to obtain the required workability on themortar mix.

Fig. 2 Cutting of trees and cleaning of the slope before pacing
the mesh (Sasagatake mountain, Iga city, Mie)
Following the shape of the slope surface, one layer of wire
mesh was placed on the slope spreading from the top tobottom of the slope. The mesh was tied by the hook or
anchored pin with the slope. The most common size of theanchored pin was 13-16 mm in diameter and 20-400 mm inlength depending on the position slope. The placement
procedure of the wire mesh is shown in Fig.3. The anchoring
pin used to fix the wire mesh on the slope is shown in Fig.4.

Fig. 3 Placement of wire mesh on the slope surface(Sasagatake mountain, Iga city, Mie)

Fig. 4 Anchored pin for fixing the mesh on the slope
Plastering: The plastering technique employed in this study
was done by spraying the mortar by a particular form of gun
which is supplied with compressed air. As part of an overall
apparatus, that includes a pump and an air compressor, whichenables spray applications to be effected remotely from themortar supply for the in-situ composite of slope surface. The
spray pattern of the gun is adjustable. Controlling the mortar
supply and air pressure helps to reduce the problems
associated with excessive spray formation and rebound of theapplied mortar. For stiff mixes, the mortar normally remains
in position after placing.
Curing: The objective of curing is to keep the mortar
saturated and to promote the hydration of cement. There areseveral methods of curing. In this research the moist curing
was followed for about one week. After that the composite
structure is cured for long time with the moisture that usuallycomes from soil.

3. Results and discussion
The sketches of various composite structures identified forslope surface protection are given in Figs.5-12. Based on thecost of the materials and labors, the cost of construction ofcomposite for slope surface protection has been estimated and376

given in Table 1. The shape the edge of a composite structureused for earth slope protection is shown in Fig.5. Both theupper and lower edges have vertical portions (flanges) of 30
cm or more depending on the field conditions. Due to thesetwo vertical portions of flanges, the edge of composite
structures would be fixed on the slope and prevent water entry
just underneath the composite layer. According to the need,however, one can make the structure without any flanges or
the flanges can be made with variable sizes. The
cross-sectional view of the composite structure showing the
mortar, mesh and the anchored pin is given in Fig.6. The
length of the anchored pin ranges from 200 to 400 mm withdiameter of 13 to 16 mm depending on the slope conditions
and position of the pin. The anchored pin is used to hold themesh on the surface of the earth slope during construction andto facilitate the composite structures to be remained on theslope during on-service. The average thickness of thecomposite structures used for slope protection is usually 6 to
7 mm. However, this may be little bit thicker due to theunevenness of the surface of the slope. A part of thecompleted composite structures for slope surface protection isshown in Fig.7. This figure also shows a portion of the meshand the application technique of the mortar on the mesh.

Fig. 5 Shape of the edge of composite for slope protection

Fig. 7 A part of completed composite for slope protection(Sasagatake mountain, Iga city, Mie)
Another item of composite that has been made for use in slope
surface protection is the composite frame structures(Figs.8-12). Traditional structures made of brick are neither
strong nor suitable for mass use in slope protection works.
Concrete made frame structures are found to be expensive
and intricacy in construction. Because of high adaptability ofcomposite to various structural forms and ease in constructionand economy; the composite technology can be applied in
making not only the complete layer on whole surface of theslope as shown in Fig.7 but also be employed in constructing
frame structures especially for large scale slope or the like.
The sketches of composite frame structure having joint at thevertical column is shown in Fig.8 and the completed structure
having joint in vertical frame is shown in Fig.9. For the caseof the joint in vertical frame, the joint is usually made at a
distance of b/2 from the nearest frame that has ground anchor
and the next ground anchor is made at a distance of b between
the two frames (Fig.8).

Fig. 8 Sketches of composite frame with vertical joint forslope protection
Fig. 6 Cross-sectional view of composite and the anchoredpin used for slope protection work

On the other hand, for the joint in horizontal frame, it isusually made at the mid-section of the frame with the distanced/2 from the center of the nearest frame as shown in Fig.10. In
this case usually, the ground anchor is provided at allintersectional points of the frame (Fig.10).

377

Fig.9 Completed composite frame with vertical joint for slope
protection (Sasagatake mountain, Iga city, Mie)

The design of the horizontal and vertical frames along withthe placement of mesh, skeletal steel and ground anchor isshown in Fig.12. Usually, the placement of skeletal steel andthe mesh in the vertical frame is done firstly then theplacement of skeletal steel and the mesh in the horizontal
frame is performed. As can be seen in this figure that themeshof the vertical frame is cut at the mid-section of the twohorizontal meshes and then bent it outside. This facilitates theease of placement of mortar and the aesthetics of thestructures. The ground anchor is provided at the position
close to the skeletal steel instead of the center of theintersection in order to clamp and hook the frame to beremained in its position during casting and on-service.
Horizontal frame

The completed structure having joint in horizontal frame isshown in Fig.11. This technique of composite structure forearth slope protection allows greeneries between the frames,
and thus conserves the environment. Therefore, thecomposite technology for the earth slope protection can betreated as the environmentally friendly technology and thecomposite can be termed as environmentally friendly
material, especially when dealing with the earth slope.

Mesh

Skeletal steel

Anchor pin

Vertical frameFig. 12 Method of processing the intersection point of thecomposite frame
The cost of different items used for the construction ofcomposite structures for earth slope protection issummarized in Table 1.
Fig. 10 Sketches of composite with horizontal joints for earth
slope protection

Fig. 11 Completed composite frame with horizontal joints forearth slope protection (Sasagatake mountain, Iga city, Mie)

Table 1 Calculation of the cost of composite structure forearth slope protection
Cost
Items
Cost (Yen)(US$)
Price of mortar per cubic
15200.0
126.67
meter (cement:sand=1:4)
Mortar per square meter
1064.08.87
with thickness 7 cm, (1cubic meter×0.7)
Wire mesh (dia.2mm, cc
230.0
1.92
opening 50mm) per square
meter
Anchor pin and spacer,
206.0
1.72
tools etc. per square meter
Mortar mixing, spraying,
3000.025.00
labors etc. per square
meter
Total cost composite
4500.037.5
structure per square meter
Note: 1US$=￥75

378(4500yen=60US$), Sept.01, 2011

The cost given in Table 1 is calculated based on the costs ofmaterials and labors. Though it has not been the prime
objective of the studies to compare the cost of composite
structures with other varieties of materials, it can, however,be emphasized that the composite structures are relatively
cheaper than those made with traditional materials liketimber, steel and such other materials.

expressed in this paper are those of the authors and do notnecessarily reflect the views of the sponsor.

5. REFERENCES

3. CONCLUSION
Composite structures that have that have been indentified andconstructed in this research study are some of the examples
that can be constructed for earth slope protection using
composite technology. Results obtained have demonstrated
that the utility and economy can both be achieved using very
simple techniques using locally available materials. It is
expected that the observations made here in this study willbring the new concept in gaining wide acceptance ofcomposite for the construction of strong, durable andcost-effective composite structures for earth slope protection.

4. ACKNOWLEDGEMENT
The author is grateful to Mr. Morimoto of Yamaichi
Construction Company Ltd. for supplying some information
during preparation of this manuscript. This research is partly
supported by the Research Grant No. 18580243 with funds
from Grants-in-Aid for Scientific Research given by theJapanese Government. The writer gratefully acknowledges
these supports. Any opinions, findings, and conclusions[1] Naaman AE, “Composite and Laminated Cementitious
Composites,” Techno Press 300. Ann Arbor, Michigan, 2000.
[2] Awal ASMA, Composite: “A unique material of
construction and its use in agriculture,” Bangladesh J, of
Agric. Engg., Vol.1, No.2, 1987, pp.35-40,[3] Hossain, MZ. and Hasegawa, T. “ A Study on Pre- andPost-Cracking Behavior of Composite Plates,” J. of
Composite,vol. 27, No. 2, 1997, pp. 127-142.
[4] Hossain, MZ and Inoue S, “ Compression Behavior andBuckling Analysis of Composite Elements Using the Finite
Element Method,” Journal of Composite, Vol.30, No.2, 2000,
pp.147-166.
[5] Hossain, MZ, Sakai, T and Paramasivam, P “ Bearing
Capacity of Thin Cementitious composites Reinforced withDifferent Types of Meshes,” Journal of Composite, Vol.36,
No.3, 2006, pp.833-848.
[6] Ohama, Y and Shirai, A “ Flexural behavior of polymer
composite with steel fibers,” J. of Composite, Vol.14, No.3,
1984, pp.205-210.
[7] Shirai, A and Ohama, Y “ Improvement in flexural
behavior and impact resistance of composite by use ofpolymers,” J. of Composite, Vol.20, No.3, 1990 pp.257-264.

379

380

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Study of Slump Testing for Hydraulic Plastic Grout
Hiroyasu ISHII1), Kenichi HORIKOSHI1) and Masaki KITAZUME2)
1) Technology Center, TAISEI Corporation, 2) Tokyo Institute of Technology



ABSTRACT: Hydraulic plastic grout, which consists of cement,water, admixtures and a plasticizing agent, has different properties in
its fresh state than other fluid construction materials such as mortars
and concretes. This brings specific benefits in civil construction,
especially in void filling and foundation work, because its plasticity
endures flow even under the influence of gravity.
In this research, the authors mainly focus on a conventional
quality test used on fluid materials in their fresh state, the slump test.
Although this is a very useful method and is used in various
evaluations of fluid materials, including hydraulic plastic grout, themeasured deformation (slump, flow) of hydraulic plastic grout isvery small due to its plastic property and small specific gravity.
This paper begins with a review of hydraulic plastic grout toindicate how it is used in practice. Then a simple modified slump test
is presented in which an additional vertical burden is applied to the
top surface of a hydraulic plastic grout sample. The slump and flow
values obtained through the method are presented and studied
analytically to obtain a typical property, the yield stress.

Keywords: Hydraulic plastic grout, Slump test, Void filling,
Yield stress
1. INTRODUCTION
Hydraulic plastic grout consists of cement, water, admixturesand plasticizing agent and has plastic properties in the fresh
state. The grout can be injected under pressure and endure
flow even under gravitational force. This makes it effective
for civil construction work such as void filling andfoundation work since it does not leave major voids. That is,
the properties of the grout are quite different from those of
conventional fluid materials like mortars and concretes [1].
(Fig. 1, 2)This type of grout has been used to fill the tail void during
tunnel boring machine (TBM) tunneling. Other recent
applications have demonstrated the effectiveness of thematerial in other areas, as described in the following section.
The use of hydraulic plastic grout is expected to expand as itcomes into use for repair work and remediation work related
to foundations.
In these various applications, past and future, different
grout plasticity may be used according to the purpose of thegrout as well as working conditions. This means that slump
tests will play a key role in on-site quality control.
Until now, the cylinder flow test (JHS A313) and the tableflow test (JIS R 5201) have been used to measure slump, just
as for mortar or concrete. The authors, however, take theview that the present use of slump tests for hydraulic plastic
grout needs further study due to the lack of consideration ofits different properties from conventional materials.

In this paper, a review of hydraulic plastic grout is given toshown how it works in actual applications. Then a simple
modified slump test is presented in which an additionalvertical burden is applied to the top surface of the hydraulicplastic grout being measured. Slump and flow values
obtained using this modified method are presented. Throughan analytical study, a characteristic property of the grout,yield stress, is estimated.
2. USE OF HYDRAULIC PLASTIC GROUT
2.1 Site operations
The components of a hydraulic plastic grout, generally
consisting of cement solution, admixture solution andplasticizing agent, are prepared and supplied separately
before mixing altogether for use. Each solution is so thin thatit can be transported through supply piping at a relatively low
pumping pressure. This is one of key benefits of hydraulicplastic grout as compared with conventional materials like
mortars and concretes, which have high viscosity and require
much higher pumping pressures.
The solutions are mixed together in the supply piping
immediately before injection. The mixed hydraulic plastic
grout gains plasticity in seconds. The level of plasticity
depends on the mix proportion, the cement content, the typeFig. 1 Hydraulic plastic grout

Fig. 2 Schematic diagram of hydraulic plastic grout during
injection

381

these voids.
Another application that takes advantage of the properties ofhydraulic plastic grout is the filling of limited volumes. This
technique has been developed for use in the filling ofabandoned underground cavities [2]. The hydraulic plastic
grout is used only to form partitions within the cavities, which
are too large and complex in shape to be filled with a fluid
material in one step. Each partition formed by placing ahydraulic plastic grout barrier is a limited region that can thenbe filled with a different fluid mixture. (Fig. 3)
Fig. 3 Schematic diagram of limited volume filling
technique applied to abandoned underground cavity [2]Placing barge
Tunnel element

Fig. 4 Birds-eye view of immersed tunnel crossing the
Bosphorus strait.
of plasticizing agent and other factors. Typical properties are
1,000–10,000 N/m2 vane shear strength and 140–160 mm ofcylinder flow value.
Injection work can take from minutes to hours depending on
the work plan. After injection, the grout is left to harden bycement hydration. Typical unconfined compressive strengthis 0.8–1 N/mm2.
The major benefit of hydraulic plastic grout is that goodfilling can be achieved without leaving major voids. This isbecause the mix proportion is intentionally prepared so thatbleeding will be negative, and also the grout undergoes
significant deformation under gravity during and after
injection.
2.2 Conventional Use
The original use of hydraulic plastic grout was to fill the tailvoid, meaning the gap around the segments, during tunnelingby boring machine. Its advantages in this application are goodfilling, early age strength, and the possibility of pumping long
distances through piping along the tunnel. A number ofdifferent mixtures and methods have so far been implemented
by construction firms in Japan.
Repair work and remediation work around foundations andretaining walls are among other typical applications. Gaps
sometimes form below or behind structures as a result ofconsolidation settlement, seismic effects and seepage
washout. Hydraulic plastic grout is very suitable for filling

2.3 Recent Applications
A recent application is in filling the space between immersed
tunnel elements and their prepared foundations for a 1.4 km
length of immersed tube tunnel crossing the Bosphorus Strait
in Turkey [3] (Fig. 4). This is one of the main structures
forming the Marmaray Project in Istanbul, which is a full
upgrading of the present commuter rail system and includes a13.4 km underground section.
As part of tunnel construction work, any space between thefoundation and the immersed tunnel element has to be filled
after each element is immersed and placed on the seabed.
Various conditions of the entire work led to hydraulic plastic
grout being chosen as the most suitable material for filling thespace under all 11 immersed elements [4].
Full scale injection tests were carried out in advance ofactual site work. These test injections were performed in thespace between a model slab and a stone foundation positioned
in a 9 m-diameter water tank. A mound with a height of 300
mm was intentionally formed on the stone foundation so as toobserve how the grout performed with such undulations. (Fig.
5)
Grout was injected over the whole area from only one
injection valve. After it had set, its surface was found to be
fully flat without major voids, as indicated in Fig. 6. It was
confirmed that one injection point could be used to inject up
to 6 m in distance.
For the actual work, injection valves were positioned along
two lines spaced 5 m apart, as illustrated in Fig. 7. Valves inboth lines were spaced at 8 m and they were offset from eachother in the two lines. This gave a valve-to-valve spacing of
6.6 m. Injection took place in order from one end of theelement, using approximately 30 valves per tunnel element.
The injection valves were used as arrival valves for grouting
from the preceding valve; arrival of the grout was mostly
confirmed by opening neighboring valves and allowing grout
to escape.
3 SLUMP TEST FOR HYDRAULIC PLASTIC GROUT
3.1 Application of Slump Test
Slump tests (or flow tests) are a well-known method of
evaluating the consistency of construction materials such asmortars, concretes, and various kinds of grout. The material
to be tested is placed in a vessel, of which several sizes andshapes are used according to sample composition or fluidity.
The cylinder flow test (JHS A 313; cylinder dia. = 80 mm,

382

Land side

Offshore side

Plane view
Unit: mm

1500

7000

3300

Immersed tunnel element
Mixer
Three supply lines for the solutions

Mound(=1,300mm)

5000

Injection valve
Injection valve

Arrival valve

Injection space
Average height=0.5m
Model slab
Sectional view

General area covered by individual injection

Arrival valve
Injection valve
Model slab

Injection space(h=500mm)

2nd
15.3m

Foundation cobble

4th

5.25m

6th

8th

10th

6.6m

Mound

1st

Fig. 5 Plane and sectional view of full-scale model test

General grout movement

Injection valve

3rd

5th
4.0m

7th

9th
Injection valve

8.0m

Fig. 7 Injection sequence for construction of immersed
tunnel foundations
Although these tests work efficiently as checks on thequality of grout, the authors raise two concerns for further
consideration:- The higher yield stress as well as the lower unit weight ofhydraulic plastic grout as compared with mortars andconcretes mean that deformation is generally poor during
flow tests.
- Although tamping 15 times as stipulated in the table flow
test effectively results in a larger deformation, the flow
value obtained is merely an index and does not directly
represent the viscous property of the grout.
In order to modify the conventional procedures to take into
account these two concerns, there are two possible
approaches that will enhance the deformation without
tamping. One is to use a larger scale and the other is to apply
an additional burden to the grout. In this study, the latterapproach is employed to examine its applicability and benefit.

Model slab

Fig. 6 Surface and section of solidified hydraulic plastic
grout after test injection
height = 80 mm) and the table flow test (JIS R 5201; cone
upper dia. = 60 mm, lower dia. = 80 mm, height = 60 mm) arein common use for measuring hydraulic plastic grout.
Although both measure the diameter of the sample material
after it has slumped, the former allows the grout to flow undergravity only, while in the latter the grout is tamped 15 times.
A target flow value of 160–180 mm was set for the mixture
developed as a limited volume fill material [2].

3.2 Theoretical Approach
After the first simple theoretical analysis relating slump toyield stress carried out by J. Murata [5], a number of studies
using similar methods have been published for mortars andconcretes as well as for mineral and industrial suspensions[6]–[9]. This study refers to an analytical model developed
for cylindrical geometry and which was generalized indimensionless form [8]. The model described below is basedin part on this conventional one, but a burden is newly added
to enhance grout deformation.
Conventional slump theory
Figure 8 is schematic diagram of the deformation and stressstate at each stage. Initially, the pressure, P|z,0, at any given
height, z, can be approximated as(1)
P  zg

383

z ,0

Fb
S1

r
dz0

Fb

z

|z,0dz|z,1dz|z,1

'|z,1

|z,1
A|z,1 , r|z,1

S2

dz|z,2

|z,2

A|z,1 , r|z,1
A|z,2 , r|z,2fa) Initial stateffb) Conventional slumpc) Applying a burden

fd) Enhanced slump

Fig. 8 Schematic diagram of deformation and stress state at each stage
where  represents specific gravity of the sample. The
maximum shear stress, |z,0, is, assuming an ideal elastic solid,
equal to half the pressure.
(2)
  0.5zg

Table 1 Mix proportion of hydraulic plastic grouts used forthe slump tests (for 1 m3)

z ,0

As illustrated in Fig. 9, the initial shear stress is distributed
linearly from 0 at the top to a maximum at the base. Toward
the top of the sample, the stress experienced may be less than
the yield stress; without yielding, no deformation takes place.
On the other hand, lower down in the sample, stress may
exceed the yield stress; flow then occurs until the stress falls
to the yield value.
(3)
A  A0  
f
z ,1

A

B

z ,0

where A0 A|z,1 represents the cross-sectional area of thecylindrical specimen in the initial state and in conventional
slump, respectively. With the height divided into thinelements of thickness dz, the volume conservation
assumption gives the following equation.
(4)
dz0  A0  dz1  ACTwice the value of r1, derived from S1, at the bottom is
specified as the flow value.
Enhanced slump theory
A burden, Fb, is applied in the conventional slump state, as
illustrated in Fig. 8(c). Based on same simplifications andassumptions used above, pressure, P’|z,1, and shear stress, '|z,1,
at any given height can be assumed as the following.
(6)
P'  P  Fb / A
z ,1

z ,1

z ,1

 ' z ,1  0.5P' z ,1

(7)

In elements where the shear stress exceeds the yield stress,
additional deformation takes place. The stress balance andvolume conservation give the following equations.
(8)A  A  '  f
z,2

z ,1

z ,1

dz z ,1  A z ,1  dz z , 2  A z , 2(9)Accordingly dz2 can be specified and the enhanced slump, S2,

M1

M2

320

240

60

60

2

2

246

309

45

35

552

516

20

20

can be numerically calculated by accumulating dz1–dz2.

S 2  S1   (dz1  dz 2 )z ,1

where dz0 and dz1 are the element thickness in the initial stateand in the conventional slump state, respectively.
Accordingly dz1 can be specified and slump, S1, can be
numerically calculated by accumulating dz0–dz1.
(5)
S1  (dz0  dz1 )Solution
Oridinal portland cement(kg)
Solution B(L)
Admixture
(L)
Water
(L)
Bentonite
(kg)
Water
(L)
Sodium Silicate
(L)

(10)

Twice the value of r2, derived from S2, at the bottom is
specified as the flow value.
3.3 Laboratory Tests
In order to investigate the actual slump value obtained by the
proposed enhanced method, slump tests were conducted withboth slump and flow values were recorded.
The two mixtures tested here were typical hydraulic plastic
grouts. Their mix proportions are summarized in Table 1.
Two tests were conducted for each mix. This allowed slump
to be measured with two different yield stresses for the same
mixture; the setting time caused cement hydration to increase
the yield stress of the grout.
The cylinder used for the slump tests had a height of 80 mm
and a diameter of 80 mm. The cylinder was first filled withthe hydraulic plastic grout. It was then lifted away, allowing
the grout to collapse under its own weight. The slump height
and flow diameter of the grout were measured as aconventional method.
In order to measure the enhanced slump, a burden wasapplied using a specially fabricated cup and small lead balls.
After the conventional measurement of slump as describedabove, these were placed on top surface of the grout to

384

100
○ M1-15min □ M1-60min● M1-15min ■ M1-60min

Slump (mm)8060Theoritical: mortar

40

Theoritical: hydraulic plastic grout

20
0
280

Flow (mm)240
Theoritical: mortar

200

Theoritical: hydraulic plastic grout

160
120
80
0

200

400
Yield stress (N/m2)

600

800

Fig.10 Relationship of yield stress against conventional
slump and flow value
280

Enhanced flow (mm)Fig. 9 The conventional slump test and the enhanced testwith a burden applied using a specially fabricated cup andsmall lead balls
Table 2 Results of conventional slump and enhanced
slump
Conventional

Enhanced

Mix.
&
Time

Flow
(mm)

Slump
(mm)

Flow
(mm)

M1
15 min.

105 x 103
Ave. 10425

122 x 113
Ave. 118

M1
60 min93 x 91
Ave. 92

16

M2
30 min.

118 x 112
Ave. 11535

M2
120 min

96 x 97
Ave. 9718○ M1-15min □ M1-60min● M2-30min ■ M2-120min

240
200

Fb=4.0 N

160

Theoritical: Fb=4.0 N
Theoritical: Fb=2.5 N
Theoritical: Fb=0 N

120
Fb=2.5 N

80

Slump
(mm)

0

Fb=2.5N

Fb=4.0N40

Fb=4.0N
142 x 142
Ave. 14248

Fb=4.0N
133 x 131
Ave. 132

400
Yield stress (N/m2)

600

800

Fig.11 Relationship of yield stress against enhanced flow
value

38

124 x 121
Ave. 123

200

43
2

Note: Fb=2.5 N and 4.0 N corresponds to 498 N/m and
796 N/m2 against the cross-sectional area of thecylindrical specimen in the initial state.
increase grout deformation. The additional slump caused by aparticular burden was measured.

3.4 Results and Discussions
Test results
Figure 9 shows hydraulic plastic grouts undergoing the twoforms of slump testing. Under enhanced deformation (lower
photo), the grout has expanded in diameter almost linearly
from top to bottom. This infers that all of the grout reachedyield in respect of the theory given in the section above. Fourtest results are summarized in Table 2, where a burden of 2.5
N was applied to grout M1 for 15 minutes and a burden of 4.0
N was applied in the remaining cases to M1 for 60 minutes, toM2 for 30 minutes, and to M2 for 120 minutes.
Theoretical calculation for conventional slump
Conventional slump and flow was first calculated
theoretically. The results are plotted in a relationship of grout
yield stress against slump and flow in Fig. 10. In addition to
the results using the actual specific gravity of 1.3, another385

calculation using a specific gravity of 2.0, corresponding totypical ordinary mortar, is shown for reference purposes inthe same figure. It is inferred from this figure that:- The conventional slump measurement is applicable tohydraulic plastic grout for yield stresses of up to
approximately 500 N/m2. For other cases, no effective
deformation occurs and slump and flow values are
identified as zero.
- This range of yield stress for the reference results extends
to 700 N/m2 due to the greater shear stresses in thespecimen resulting from its higher specific gravity.
This relationship between yield stress and conventional
slump and flow value was then used to identify the yieldstress of the tested specimen, as indicated in the diagram by
the arrows. Clear differences are found between the estimatedyield stress values given by the slump and flow values. The
reason for this cannot be explained and is to be examined infurther work.
Theoretical calculation for enhanced slump
The relationships between yield stress and the enhanced flow
value are plotted in Fig. 11 based on the theory described in
the section above. One line in the figure represents the resultwith a burden of zero, which is equivalent to the conventional
slump test. The others are results for burdens of 2.5 N and 4.0
N, corresponding to the tests conditions. It can be inferredfrom this figure that, whereas the conventional slump test isapplicable up to a yield stress of 500 N/m2, this cantheoretically be extended to as much as 800 N/m2 with theenhanced test. The measured flow values summarized inTable 2 are plotted in the same diagram, where the estimatedyield stress in Fig. 10 was employed. The plot with Fb = 4.0 N
generally matches the calculated results for the corresponding
burden. The enhanced flow caused by the Fb = 2.5 N burdenlies below this curve, and matches the theoretical results well.
This indicates that the diagram can be used just as Fig. 10 was,to evaluate the yield stress from flow values.
Although further verification is preferred though tests on
different mixtures of hydraulic plastic grout, this study
suggests that the proposed enhanced slump test is an effective
way to estimate the viscosity of hydraulic plastic grouts more
precisely and over a wider range than the conventional slump
test.

5 REFERENCES
[1]

[2]

[3]

[4]

[5]
[6]

[7]

[8]

[9]

4 CONCLUSION AND REMARKSThis paper first summarizes the properties and uses ofhydraulic plastic grout, and then proposes an enhanced slump
test that provides better estimates for this material, which
differs from conventional materials to which slump tests areapplied, like mortars and concretes.
Hydraulic plastic grout is a material particularly suited for
repair and remediation work. Usage may be expected toincrease as the proportion of such types of work increases incomparison to new construction. The authors aim to carry out
further studies of this beneficial material, looking into its
properties, methods of estimating them, and its site
application.
386

Miki G., Uechi H. and Shimoda, K., “New backfill material and its
grouting system,” Journal of the Society of Materials Science, Japan,
Vol.31, No.341, 1982, pp.138-143. (in Japanese)
Sakamoto A. et al., “Development and applications of limited area
filling techniques for abandoned underground cavities”, Journal ofJapanese Society of Civil Engineer, Vol.62, No.3, 2006, pp. 546-557.
(in Japanese)
Ingerslev L.C.F.et al., “Marmaray project: requirements for the designand construction of the Bosphorus tunnel,” Proc. of International
Tunneling Association (ITA) World Tunnel Congress, Istanbul, 2005,
pp. 1269-1275.
Ishii H. et al., “Construction of immersed tunnel foundations using
hydraulic plastic grout”, Proc. of 14th Asian Regional Conference on
Soil Mechanics and Foundation Engineering, May 2011, CD-ROM.
Murata J., “Flow and deformation of fresh concrete”, Materiaux etConstruction, 1984, pp.117-129.
Aaron W. Saak et al., “A generalized approach for the determination ofyield stress by slump and slump flow”, Cement and Concrete Research,
No.34, 2004, pp.363-371.
S. Clayton et al., “Analysis of the slump test for on-site yield stressmeasurement of mineral suspensions”, Journal of Mineral Processing,Vol.70, 2003, pp. 3-21.
N. Pashias and D. V. Boger, “A fifty cent rheometer for yield stressmeasurement”, Journal of Rheology, Vol.40, No.3, 1996, pp.
1179-1189.
W. R. Schowalter and G. Christensen, “Toward a rationalization of theslump test for fresh concrete: Comparisons of calculations andexperiments”, Journal of Rheology, Vol. 42, No. 4, 1998, pp. 865-870.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Behavioral Comparison of Coarse Aggregate from Various Quarry Sites ofPakistan, Effect on Hardened Concrete, Economics and Environment
Faisal Malik 1, Dr. S. Muhammad Jamil PhD 2
1

Graduate Researcher, School of Civil and Environmental Engineering, National University of Science and Technology,
Islamabad, Pakistan.
2
HOD, Department of Geotechnical Engineering, National University of Science and Technology, Islamabad, Pakistan.
ABSTRACT: Coarse aggregate quarry sites are available indifferent regions of Pakistan. Lack of geotechnical data ofthese sites has led to a biased practice of using crushed stoneeither from “Margallah” or “Sargodha”, located in the North ofthe country. Common practice of using these quarry sites isleading to enormous transportation cost, for projects towards
South of the country. This practice adversely affects the project
economics. Seven coarse aggregate quarry sites were identified,in various parts of the country. The material from each site wastested for thirteen different tests as per ASTM/BS to include
Specific Gravity and Absorption, Shape, ChemicalComposition through X-Ray Fluoresce (XRF), Microfines,
Characterization of Microfines through X-Ray Diffraction(XRD), Clay lumps and friable Particles in Aggregate, Loss
Angeles (LA) Abrasion, Aggregate Crushing Value, Aggregate
Impact, Sodium Suphate Soundness and Compressive Strengthof Concrete. It revealed that basic constituent of six quarrysites was lime stone (sedimentary rock) against Sargodha
quarry site where it is igneous, and show indifference
properties. The analysis led to 60% cost effect solution towards
the South. The study also generated comprehensive databasefor the construction industry. Over reliance on existing known
sources will be reduced, which is depleting the Magallah hills
resulting into big environmental issue.
Keywords: Coarse aggregate, Concrete, Pakistan, Margallah,
Quarry sites

1. INTRODUCTION
Since up to approximately 80 percent of the total volume
of concrete consists of aggregate, aggregatecharacteristics significantly affect the performance offresh and hardened concrete and have an impact on thecost effectiveness of concrete [1]. Relative density is notnecessarily related to aggregate behavior. However, it
has been found that some aggregates compounds of shale,sandstone, and chert that have somewhat low specificgravity may display poor performance, particularly in
exposed concrete in northern climates [2]. Aggregate
porosity (absorption) may affect durability as freezing ofwater in pores in aggregate particles can cause surface
popouts [3]. Aggregate characteristics of shape, texture,and grading influence workability, finishability, bleeding,pumpability, and segregation of fresh concrete and affectstrength,
stiffness,shrinkage,
creep,
density,permeability, and durability of hardened concrete [4].
Shape is related to three different characteristics:
sphericity, form, and roundness [5]. Research shows that
there is a clear relationship between shape, texture, andgrading of aggregates and the voids content of
aggregates [6]. Coatings, the layers of material covering
the surface of aggregate, can increase the demand forwater and can impair the bond between paste andparticles. Sometimes these coatings are formed bymaterials that can interact chemically with cement,
negatively affecting concrete [7]. Soundness, according
to Forster [8], is the aggregate resistance to weatheringthat primarily includes resistance to freezing andthawing, and to a lesser extent, resistance to wetting anddrying; and heating and cooling. A sound aggregate hasa satisfactory durability factor when used in properly
mature concrete with enough air void content [9].
Pakistan has coarse aggregate quarry sites through its
length and breadth. However proper developed quarrysites are restricted in Northern or central region of thecountry which are contributing the construction activities
in the country. Detailed characterization of coarse
aggregate even for the northern quarry sites is notavailable. Comprehensive testing mechanism was This
study has not only yielded coarse aggregatecharacterization database as per ASTM/BS but also
identified 3 new quarry sites towards the West and Southof the country. Identification of three quarry sites has
resulted in cheaper coarse aggregate in South of country
and lowering reliance on the Northern quarry sites thus
slowing environmental issues linked to depleting
Northern quarry sites.
2. IDENTIFICATION OF COARSE AGGREGATEQUARRY SITES
Emphasis was laid on identification of new coarse
aggregate quarry site towards the South of the country,
evaluating its material and engineering properties also
drawing comparison with existing quarry sites. The
locations of under study coarse aggregate deposits are
shown
in
Fig
1.

387

Fig 1. Locations of Under Study Coarse Aggregate Deposits
3. EXPERIMENTAL PROGRAM
3.1 Specific Gravity and Absorption
The test was to ascertain specific gravity and percent
absorption to for material characterization. The test was
conducted as per ASTM C 127-88[10]. The results areshown in Fig 2.
2.62 2.73 2.61 2.65 2.67 2.67 2.65

% Element

3.00

3.3 Mineral Structure through X ray Florescent(XRF)The test was selected to indentify basic material
constituents of coarse aggregate. It was conducted as per
ASTM C 295 [12] and revealed indifference value for six
sites indicating lime stone. The results are in Fig 4.

2.00
1.00

0.70 0.55 0.73 0.73 0.51 0.64 0.56

0.00

Specific Gravity

100
80
60
32.1942
23.5286
40 24.9634
4.7967 8.8212
20
0
Ca
Fe
Si
Mg
Al
Margallah
DM Jamali
RR Door% Absorption

Sargodha
Karachi

D.G.KhanQuettaFig 2. Specific Gravity and Absorption

Fig 4. Mineral Structure through XRF

3.2 Shape
The test was conducted as per BS 812 105.1/2 1998[11].
The results are shown in Fig 3.

3.4 Percentage of Microfines
The test was conducted as per ASTM C 117-95 [13].
The results are shown in Fig 5.

14.95 13.90 12.69 14.13 12.48

11.10% Microfines

10.89

11.58 8.34 8.41 8.15 8.28
6.82
5.59

Flakiness Index

0.76
0.42

0.42
0.20

Elongation Index

Fig 3. Elongation and Flakiness Index

Fig 5. Percentage of Microfines

20.58 0.54
0.15

% ACV

27.0
18.49

23.82 22.97 22.82 22.64

19.62

Fig 8. Aggregate Crushing Value (ACV)
3.9 Aggregate Impact Value (AIV)The test was aimed at identifying material behavior
against impact. The test was conducted as per BS 812112 [17]. The results are shown in Fig 9.

% AIV

17.82

17.30
14.41
11.30

13.46 13.01
12.53

0.59

0.10

0.07

0.17

0.10

0.00

0.10
Fig 9. Aggregate Impact Value (AIV)

Fig 6. Percentage of Clay Lumps and Friable Particles
3.7 Abrasion and Impact in the LA Machine
The test was aimed at finding the material strength andconducted as per ASTM C 131-96 [15]. The results areshown in Fig 7.
24.8% Abrasion

The test was aimed at identifying material behavior
against external forces. The test was conducted as per BS
812-110 [16]. The results are shown in Fig 8.

24.7
20.77
15.4

3.9 Soundness of Aggregate by Use of Sodium
Sulphate
The test was aimed to find the material strength against
chemical reaction. The test was conducted as per ASTM
C 88-90 [18]. The results are shown in Fig 10.
% Loss due sodium
Sulphate% Clay lumps or friable
particles

3.5 Characteristics of Microfines through X-Ray
Diffraction (XRD)
The test aimed at microfines characterization to know if
any quarry site contain deleterious microfine or otherwise.
The results are shown table 1. The tests result of XRD,
given in table shows, indifference value for six sites
indicating crushed stone dust i.e. no undesired microfines.
Table 1: Characteristics of Microfines
Site
CompoundCalcium Carbonate, Ca C O3,Margallah
Calcite, synSargodha
Calcium, Ca, beta - phase
Calcium Carbonate, Ca C O3,
DG KhanCalcite, syn
Calcium Carbonate, Ca C O3,
DM Jamali
Calcite, syn
Calcium Carbonate, Ca C O3,
Nooriabad Karachi
Calcite, syn
Calcium Carbonate, Ca C O3,
Sohrab Quetta
Calcite, syn
Calcium Carbonate, Ca C O3,
RR DoarCalcite, syn
3.6 Clay Lumps and Friable Particles in Aggregates
The test was selected to know if any deleterious material
attached with coarse aggregate from any quarry site,which minute and with the allowable limit of ASTM. The
test was conducted as per ASTM C 142-78 [14]. The
results are shown in Fig 6.

24.1

18.0

16.27

9.0

9.0

9.0

10.0 10.0

5.0

6.0

Fig 10. Soundness of Aggregate by Use of Sodium
Sulphate
Fig 7. Abrasion & Impact in LA Machine

3.10 Compressive Strength of Concrete

3.8 Aggregate Crushing Value (ACV)

3

The concrete mixes were prepared using coarse aggregate
from seven different sources and remaining variables asconstant with the cement: sand: coarse aggregate ratio of
1: 2: 4 (by weight) respectively, cement content (201.7kg/cum) and water cement ratio of 0.5. The course andfine aggregates were used in SSD condition. The fine237

Compressive Strength (Kg/sq cm)

250
212

212

212

187

200
150

211

147

145

78

74

156

222
203

216

211

185

183

166

100
100

aggregate used was taken from lawncepur quarry locatedin the Northern region. The cement used was Fauji
cement (type 1, Portland cement). The water used was
normal tap water without any undesired colour. The test
was conducted as per ASTM C 39-96 [19]. The results
are shown in Fig 11.

185

182

159

152

86

90

8186

93

DM Jamali

Karachi

QuettaRR Haro14 Days

28 Days500Margallah

Sargodha

DG Khan

3 Days

7 DaysFig 11. Compressive Strength of Concrete
4.

TEST RESULTS AND DISCUSSIONS

4.1 Petrographic Analysis
The test led to conclusion, that out of seven coarse
aggregate quarry sites, six i.e. (Maragllah, DG Khan, DM
Jamali, Nooriabad Karaachi, Quetta and River Run River
Doar Abbottabad), limestone as basic rock constituentsand the seventh site (Sargodha) has Basalt as basic rock.
Both rock types exhibit satisfactory material, engineering
behavior, thus can be used for construction purposes.

Compressive strength of concrete based on the aggregate
from quarry sites also indicates that the strength andbehavior is quite satisfactory.
4.5 The Result
Series of test result show that three minimal exploited
site towards the west of country can be used withsignificant degree of confidence in construction industry,
reducing dependence on depleting coarse aggregate
source.
5. ECONOMICAL IMPACT

4.2 Basic Material Characteristics
Test results of specific gravity, absorption, microfinesand clay lumps support each other, and follow laid down
limits. It also indicate, specific gravity as basic property
if an aggregate has higher values for it (within allowable
limits), the aggregate behavior will be better.
4.3 Material Strength Characteristics
Results of Loss Angeles abrasion test, Aggregate
Crushing Value and Aggregate Impact Value tests follow
same trend and support one another results, indicating
sufficient physical strength for aggregates underconsideration. Depending upon project importance, any
one of three test may be conducted that will lead tostrength parameters.
4.4 Concrete Compressive Strength

5.1 Demand and Supply
Pakistan is self-sufficient in primary aggregates and there
are significant regional imbalances in consumption,
which require balance in inter-Pakistan movements ofaggregates. Present reliance/share of aggregate quarrysites is given Fig 1. Consequently there is substantial andincreasing movement of aggregate in the Pakistan
straining the economy and raising construction cost.
5.2 Haulage vs. Cost
Transporting aggregates, and the effects of aggregatestransport, aggregate products are generally used withhundreds of kilometers in Pakistan as compared to UK
where it is within a thirty-mile radius of the quarry fromwhich they have been extracted. One clear example ofhaulage is from D.G Khan to interior/South Sind,4

involving one way distance of 400 - 800 kms withapproximate haulage cost of Rs (6000 – 18000)/400 cft.
Cost Effect is reflected in table 2.
Table 2. Cost Effect of Transportation
Material
Cost of
Crush

TransportationCost(Rs/Cum)(Rs/Cum)

DG Khan

420

470

890

Sukhur

420

1325

1745

Hyderabad

420

1782

2202

Region ofUse

Total Costof Crush(Rs/Cum)

5.3 Cost EffectivenessCoarse aggregate cost comparison from the existingarrangement to the proposed arrangement is upto 60%
cost effective. Comparison is given in table 3.

Fig 12. Recommended Regions of Supply

Table 3. Cost Effectiveness as result of studyRegion/placeof UseSukhurHyderabad

Cost of Local
Crush(Rs/Cum)
890 ( DM
Jamali)
890(Nooriabad)

Cost of
DG
Khan
Crush
(Rs/cum)

%age
Decre
ase

1745

49 %

2202

60 %

5.3 Recommended Regions of Supply
The experimental study of the sites towards the south ofPakistan has revealed that there is no significant
difference in crush from Margallah / Sargodha / DG
Khan and Nooriabad Karachi / DM Jamali / Quetta.
Based on study three new quarry sites were indentified
with acceptable range of material properties. Hencecoarse aggregate from the later sites can be used withsame confidence for engineering purposes. Therefore
redistribution of supply regions recommended speciallyeffecting towards the South of the country. This will also
reduce the project cost considerably. New suggested
areas of coarse aggregate supply from particular quarrysite are given in Fig 12.

5.4 Benefits Visualized




6.

Reduction in cost of construction by minimizing thetransportation cost of crush.
Minimizing the dependence of construction projects
on a specific source of crush.
Development of local economy.
Indirect Cost savings from vehicle/roads wear andtear.
ENVIRONMENTAL ISSUE

6.1 The Issue
Margallah hills coarse aggregate quarry site mainlycomprises of sandstone and shale, which is very suitablefor construction of building and bridges. It is the main
supplier of limestone for the construction, and wellknown/exploited in construction industry. Over relianceon one quarry site for coarse aggregate is/will lead toenvironmental issue. Margallah is point incase. The issue
is due to effects of Mining and crushing Margallah hills.
That is why the beautiful Margallah hills are being eaten
away causing large scale environmental degradation andair pollution. Based on the interpretation of the aerial
photographs and studying other relevant data, it can beconcluded that if the present quarrying remain continue, verysoon the western side of the Margalla hills will be diminishes[20].

6.2 Envisaged effect of study to the issue
The study led to identification of three new sources ofcoarse aggregate towards the South of country. As andwhen these sources are exploited, the trickledown effectwill relieve Margallah, Sargodha and DG Khan, fromever increasing mining activity. Probability study
indicates that 20-30% decrease in demand from
Margallah, Sargodha and DG Khan, resultantly thedecreased mining activity. Although additional measure
will be required to avoid the environmental damage, yet
this study will help in delaying the issue.

5

7

CONCLUDING REMARKS

Based on laboratory investigations and data analysis,
following conclusions are drawn: Total seven coarse aggregate sites were selected for
aggregate characterization, to include Maragllah,
Sargodha, DG Khan, and River Run River Doar
Abbottabad towards the north and center while DM
Jamali, Nooriabad Karachi, and Quetta were selected
towards south of Pakistan. This will allow
development of three new coarse aggregate quarrysites towards South of the country.
 Aggregatecharacterization
procedure
devised/adopted has given details of physical,
engineering and petrographic properties of coarse
aggregate thus all inclusive database has beencreated.
 Out of seven coarse aggregate quarry sites, six i.e.
(Maragllah, DG Khan, DM Jamali, Nooriabad
Karaachi, Quetta and River Run River Doar
Abbottabad), limestone as basic rock constituentsand the seventh site (Sargodha) has Basalt as basic
rock. Both rock types exhibit satisfactory material,
engineering behavior, thus can be used forconstruction purposes.
 Coarse aggregate cost comparison from the existingarrangement to the proposed arrangement is upto
60% cost effective.
 Exploiting new sources in the light of study will
result, 20-30% decrease in demand from Margallah,
Sargodha and DG Khan, resultantly the decreased
mining activity – environmental hazard.
REFERENCES[1] Hudson, B. “Modification to the Fine Aggregate
Angularity Test,” Proceedings, Seventh Annual
International Center for Aggregates Research
Symposium, Austin, TX, 1999.
[2] Legg, F.E. Jr., Aggregates, Chapter 2, Concrete
Construction Handbook, ed. Dobrowolski, J.
McGraw-Hill, 1998
[3] Helmuth, R.A., “The Nature of Concrete,” ASTM
SpecialTechnical
Publication
No.
169C,
Philadelphia, 1994
[4] Lafrenz, J.L., “Aggregate Grading Control for PCC
Pavements: Improving Constructability of Concrete
Pavements by Assuring Consistency of Mixes,”
Proceedings, Fifth Annual International Center for
Aggregates Research Symposium, Austin, Texas,
1997.
[5] Galloway, J. E. Jr., “Grading, Shape, and Surface
Properties,” ASTM Special Technical Publication
No. 169C, Philadelphia, 1994, pp. 401-410.
[6] De Larrard, F., “Concrete Mixture Proportioning: A
Scientific Approach,” London, 1999.
[7] Mather, B., “Shape, Surface Texture, and Coatings,”
ASTM Special Technical Publication No. 169A,
Philadelphia, 1966

[8] Forster, S.W., “Soundness, Deleterious Substances,
and Coatings,” ASTM Special Technical Publication
No. 169C, Philadelphia, 1994.
[9] Mather, B., “Durability: Freezing and Thawing,”
Proceedings, Seventh Annual International Center
for Aggregates Research Symposium, Austin, Texas,
1999.
[10] ASTM C 29 Test Method for Bulk Density (“Unit
Weight”) and Voids in ggregate, Philadelphia, PA:
American Society for Testing and Materials, 1997.
[11] British Standards Institution, “Methods fordetermination of particle shape," BS-812-105: Part 1
and 2, 1989
[12] ASTM C 295, Standard Guide for Petrographic
Examination of Aggregates for Concrete,
Philadelphia, PA: American Society for Testing andMaterials, 2003[13] ASTM C 117, Standard Test Method for Materials
Finer than 75 μm (No. 200) Sieve in Mineral
Aggregates by Washing, Philadelphia, PA: American
Society for Testing and Materials, 1995.
[14] ASTM C 142-78, Standard Test Method for clay
lumps and friable particles in aggregate, Philadelphia,
PA: American Society for Testing and Materials,
1984[15] ASTM C 131-96, Standard Test Method for
degradation of small size coarse aggregate, byabrasion and impact in the Loss Angelus Machine,
Philadelphia, PA: American Society for Testing andMaterials, 1984[16] British Standards Institution, “Methods fordetermination of aggregate crushing value (ACV),"
BS-812-110, 1990.
[17] British Standards Institution, “Methods fordetermination of aggregate impact value (AIV)," BS812-112, 1990
[18] ASTM C-88-90, Standard test method for soundnessof aggregate by use of sodium sulphate ormagnesium sulphate, Philadelphia, PA: American
Society for Testing and Materials, 1990.
[19] ASTM C-39-96, Standard test method forcompressive strength of cylindrical concrete
specimen, Philadelphia, PA: American Society for
Testing and Materials, 1996.
[20] The Effect of Mining on Geomorphology (Detectionof Changes by Using Remote Sensing Techniques)
by: Falak Nawaz

6

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: Article Number 148

Methods for execution of concrete piles in corrosive and destructive marine
environments, based on study of the Persian Gulf marine installations
Nasser ArafatiA. Abdol Hossein Fazlib. Seyed Mehdi Mousavic. Department of civil engineering, Tafresh
University, Tafresh, Iran

A: Associate professor. b&c: Research Student

ABSTRACT: Marine environments include ions and corrosive
gases and are the place of marine living things, which are harmful
for construction materials. Hydrostatic pressures and hightemperature changes, which are current in most of the coastal areas
and marine structures, are able to accelerate destruction of
construction materials. Marine structures and port installations
existing in ports are considered as important infrastructural
installations that due to their concrete or metal nature are exposed todecay and destruction arising from being in marine environments.
Concrete piles, which are executed in sea beds or in coasts, areaffected by corrosive factors and the factors decreasing stability.
Therefore methods and materials should be used to confront thesefactors. Undoubtedly Persian Gulf is one of the most corrosive andrisky environments of the world for execution of concrete structuresand concrete piles.
This research tries to investigate the optimum methods for executionof concrete piles and also concrete technology of these piles in
Persian Gulf in details. In order to access this aim the geotechnicalexperiences of the piles and foundations executed in Persian Gulf
were used and finally some proposals for execution of such
structures in destructive environments without any problem were
offered.

Keywords: marine environment, displacement concrete piles,
Persian Gulf, pile concrete technology, driven piles, Concrete
Materials
1. INTRODUCTION
Generally, ease of use of driven piles in marine environments
caused their more extensive application as compared withdrilled shafts. Using driven piles in marine environments wasstarted since Neolithic times. The first residents of the globe
used to utilize such piles for construction of their houses onwater in flood plains and different types of bridges on rivers.
For execution of driven piles different methods have beenused since that time up to now [1]. In the past very simple andtime-consuming methods and equipment were used but
nowadays precise and very quick and modern methods andequipment are employed.
Modern equipment can be employed by contractors to ensure
safe, quick and adequate installation of driven pipes. Global
Positioning Systems (GPS) aid in survey and accurate layoutof pile locations. New hammers are capable of energy

adjustments and monitoring to provide the required hammer
performance while protecting the piles from damage.
Instruments can be used during testing and installation toconfirm the design and pile load capabilities.
In addition to the problems related to common piles, piles
problems in marine environments include the problems
arising from the corrosive water environment. Some of these
problems are:Liquefaction of saturated sands during pile installation,
settlement around piles, heave, buoyancy and hydrostaticpressure [1],[2].
One of the most important problems we encounter in marine
environments is splash zone effect and a more important
problem is the severe corrosive effect of some of the marine
environments on piles.
At present one of the most corrosive marine environments of
the world is Persian Gulf in the south of Iran. Because of theabundant amount of salts and ions existing in thisenvironment, concrete piles are under severe decay anddestruction.
Very corrosive climatic conditions, insufficient knowledge ofthe executive personnel, weakness in construction stages,
application of improper and low quality materials, etc. aresome of the factors of different types of destruction of
concrete structures in Persian Gulf marine environments.
Current regulations of concrete production and piles
execution, which have been considered for execution in theother points of the world, are not suitable for such anenvironment and create a lot of problems. There are a lot ofsamples of problems occurred in such structures in 1990s.

1.1 general descriptions of exposure conditions
Persian Gulf is one of the most corrosive and aggressive
environments for concrete structures particularly concrete
piles; drilled piles and driven piles. As it is shown in ''Table.1''
In comparison with other seas in the world, Persian Gulf has
more chloride ions than others. This content surely increases
durability risk in these environments, also jeopardize
reinforcements to corrosion. Temperature difference between
night and day is about 25-35 centigrade,''Fig.1'' and humidity
difference is high too,''Fig.2''. [3, 4]Thus executing method and concrete technology of piles andother concrete structures is unique in this Gulf.

393

Quality control, efficient increased strength and durability,
sufficient curing, manufacturing precision are advantages of
pre stressed and pre tensioned driven piles in comparison withdrilled piles in these environments.

Table 2: ions content in world seas.

2. Backgrounds
Reinforcement's corrosion is usually most considerable issue
in deteriorated marine concrete structures, induced chloride
ion affects on durability of concrete structures in severe
exposures. Chloride ions which penetrate through cracks orvoids or diffusion through matrix cause to onset of pitting
corrosion when their concentration at the steel surface
exceeds a critical threshold value. This threshold depends on
several factors, such as pH of concrete, steel potential andpresence of micro voids in the cement paste [6, 7, 8, and 9].
One procedure to prevent or delay the corrosion is coveringpiles with epoxy resins [10], since piles are more risky intidal/splash zones, epoxy resins can protect piles fromconnection between seawater and concrete layer.
Another way is providing cathodic prevention by means ofsacrificial anodes [6],[11]. Often cathodic protection cost is
higher than strengthed concrete which has optimum strengthand durability, so high cost of this method is one of thedisadvantages of this method. Contractors prefer to increasematerial quality and quality control instead of using cathodic
protection.
Chloride ions influx through cracks that exist on concrete
cover of piles. After considerable time, corrosion starts whenions connected to reinforcements, thus preventing penetrationof chloride ions is important step in piles and marine concrete
structures. In addition chloride penetration increases voids[18] thus deterioration risk arises consequently.

Fig.1.Montly relative Humidity averages

Fig.2.Montly Temperature

ions
K++

Balticsea
180

Atlantic
sea
430

North
sea
400

Mediterraneansea
420

Persian
Gulf
450

Ca++

190

410

430

470

430

Mg++

600

1500

1330

1780

1460

So4-

1250

2540

2780

3060

2720

Na+

4980

9950

11050

11560

12400

Cl-

8960

17830

19890

21380

21450

Salt
Content

16200

32600

35900

38700

38900

Table 1: ions content in world seas.

1.2 Pile selection:
Due to high corrosion risk for steel piles, pre stressed or pretensioned concrete piles are recommended for thisenvironments [5]
Fig 3

394

Fig.3, ''illustrate the electrochemical process of steel
corrosion in moist and permeable concrete .The galvanic cell
constitutes an anode process and a cathode process. The
anode process cannot occur until the protective or the passiveiron oxide film is either removed in an acidic environment(e.g., carbonation of concrete) or made permeable by theaction of CL- ions. The cathodic process cannot occur until asufficient supply of oxygen and water is available at steel
surface. The electrical resistivity of concrete is also in thepresence of moisture and salt'', Mehta et al [11]Fig.5: Diagrammatic representation of reinforced concrete
cylinder exposed to seawater (From Mehta, P.P.,
Performance of Concrete in Marine Environment, ACI SP-65,
pp.1_20, 1980) [23].
"The type and severity of attack on a concrete sea structure
depends on the conditions of exposure. The sections ofstructure that remain fully submerged are rarely subjected to
frost action or corrosion of embedded steel. Concrete at thisexposure condition will be susceptible to chemical attacks.
The selection above the high-tide mark will be vulnerable toboth frost action and corrosion of the embedded steel. The
most severe deterioration is likely to take place in the tidal
zone because here the structure is exposed to all kinds ofphysical and chemical attacks" Mehta et al [11].
3. Materials

Fig.4

Fig.4, ''shows that depending on the oxidation state, thecorrosion of metallic iron can result in up to six times
increase in the solid volume'', Mehta et al [11]Fig. 5

3.1 Cement and replacements
There was a view that cements type (V) is suitable andalso more durable than type(I) ,(II).but experiences in
concrete structures and concrete piles in Persian Gulf showsthat structures which are made of cement type (V) has moredeterioration than structures which are made of cement type(I),(II) [1].Also M. Naderi showed that concrete prisms whichcontains cement type( I ) has more chloride penetration
resistance than concrete prism which contains cement type(II) [10]. Using type (V) reduces durability of concreteconsiderably. Moreover references [13],[14] prove this claimthat replacing fly ash by 15% cement content increase
strength and durability of concrete structures and moreover itis proven that it can increase initiation time to corrosion,15%
fly ash and w/b=0.4 can delay corrosion initiation until 60
years [15].
As reference [16] shows, replacing calcium chloride by 2-4%
cement content , that is used as setting accelerator wasrecognized as a corrosion cause in pre stressed tendons. So it
is prohibited to replacing
calcium chloride as cement
replacement or hardening accelerator.
Although voids and entrained air can increase permeability
and consequently decrease durability but according Haynes,
H. H, air entrain by 2-6 % doesn't increase permeability[15],and it is recommended for concrete piles in Persian Gulf. It
can reduce cracks which are caused by temperature gradientsand also it can increase workability and slump (consequently
it will be easier to compacting and casting successfully).
According to A.Shayan experiments, replacing Glass
powder as a pozzolanic material reduces chloride ion
penetration [19] utilization this material in concrete is highly
recommended in casting marine piles and marine structures.
Coastal concrete structures in Netherland that made bycements with blast furnace slag showed no corrosion after 30
years. Although it is possible that corrosion initiated after 20
years but no corrosion detected in 88% of 64 structures [21]3.2 Aggregate395Aggregates contain 60-80 % concrete volume and areconsidered as inactive material in concrete structures but
it is not allowed to neglect effect of aggregate types on
marine structures especially which are in severe exposureand aggressive environments. Aggregates are cause ofabnormal dimensional stability and durability depletion.
A.Shayan experiments, which is a comparison between
non-reactive, slowly reactive, highly reactive behavior inseawater proves this claim that it is not allowed to utilize
highly reactive aggregates and it is suggested not to use
slowly reactive aggregates in high temperature[20].Expansion of highly reactive is recognized asdeterioration cause in marine concrete structures in southof Iran and Persian Gulf shores.
Limestone and non-reactive dolomite aggregates or
sufficiently high level of other forms of carbonate
calcium
have
desirable
corrosion
properties
R.E.Melchers et al [10]. Limestone favorable corrosion
properties can observe in Escambia Bay Bridge, piles intidal zones didn’t have any significant corrosion after 30
years [20].
It is highly recommended to rinse aggregate before use. It
lets concrete to reduce alkalinity and consequently
reduce expansion of pile thus it will be easier to reducesurface cracks.
3.3 Water Quality
Due to delicacy of piles, it is not allowed to use sea waterfor producing concrete, although concrete prism which ismade by sea water can reach up to 90% of compression
strength of ordinary water (edible), but it isn’t recommended
to use seawater or any salty water.
4. Execution and design notice
Reducing w/b is the most important element to have
durable, strengthen, corrosion protected, and any otherdesirable issues. As Woosuk [15] showed, lowering w/b
about 0.05 is more effective than 13mm (1/2 inch)
increment of cover thickness. It means that the piles with
51mm(2inch)clear cover and w/b=0.35 act better than thepiles with 76mm(3inch) cover and w/b=0.4 .In marine
splash/tidal exposures, Woosuk et al [15], showed that
service life of piles with w/b=0.3 concrete mixture and51mm (2 inch) clear cover is more than 100 years.
Thermal gradient during steam curing, transporting by
inexperienced performers, using improper driving
hammer, inadequate driving, should be controlled byexecutors to prevent premature cracks.
Protecting pile by covering epoxy resin that is a method
to prevent of corrosion needs highly experienced workers
to isolate piles completely from moisture.
5. Conclusion
As it was observed, with due attention to the very especial
conditions of the Persian Gulf with a view to both very

corrosive environment and abundant number of coastal
constructions it was tried to present the best method for piles
selection and execution with use of the most recent researches
and executed samples. In brief the results were as follows:-Use of pre-stressed or pre-tensioned driven piles-Covering piles with epoxy resins for splash zones-Using cement type (1) or (2)-Replacing fly ash by 15% cement or even higher-Not using calcium chloride-Aid entrain by 2-6%
-Using glass powder as a pozzolanic material-Using non-reactive aggregates-Using limestone and non-reactive dolomite aggregates ofcarbonate calcium-Rinsing aggregate before use-Not using marine water for concrete production-Reducing w/b-Employment of experienced and trained personnel
Using the above-mentioned method causes much increasein life of the piles executed in the Persian Gulf environment,in such a way that with execution of these methods corrosion
is started much later. In such conditions we can make sure of
the confident performance of the executed structure during its
lifetime.
In addition to increment of piles execution quality, anotherpreference of execution of the above-mentioned methods is
economizing of construction and execution activities of such
resistant to corrosion piles, which is another positive factor
for encouragement of contractors for execution of thesemethods.
6. REFERENCES[1]Foundatioun Engineering:In the wet design and counstruction of civil
works projects,U.S Army corps of engineers,No:1110-2-565[2] R.J Jardine, Review of technical issues relating to foundations andgeotechnics for offshore installations in the UKCS, First published2009,Imperial college London[3] Concrete in Hot countries, office of the deputy for technical affairs
bureau of technical affairs and standards, No: 184[4] A.R. A1-Rabiah, Rasheeduzzafar, Roger Baggott, Durability
Requirements for Reinforced Concrete
Construction in Aggressive Marine Environments, Marine Structures 3(1990) 285-300[5] Joseph e. Bowles, Foundation analysis and design, fifth edition, Mc-Graw
hill[6] Luca Bertolini, Elena Redaelli, Throwing power of cathodic prevention
applied by means of sacrificial Anodes to partially submerged marine
reinforced concrete piles:
Results of numerical simulations[7] L. Bertolini, B. Elsener, P. Pedeferri, R. Polder, Corrosion of Steel inConcrete.
Prevention, Diagnosis, Repair , Wiley-VCH, Weinheim, 2004.
[8] C.L. Page, Corrosion and protection of reinforcing steel in concrete, in:C.L.
Page, M.M. Page (Eds.), Durability of Concrete and Cement Composites,
Woodhead Publishing Ltd., Cambridge, 2007, pp. 156–186.
[9] A. Bentur, S. Diamond, N. Berke, Steel Corrosion in Concrete:
Fundamental and Civil
Engineering Practice, E&FN Spon, London, 1997.
[10] R.E. Melchers, C.Q. Li , Reinforcement corrosion initiation andactivation times in concrete structures exposed to severe marine
environments, Cement and Concrete Research[11] p.k.mehta,p.j.m.monteiro, concrete microstructure, prosperities andmaterial, chapter5396

[12] M. Naderi and S. A. K. Mousavi, Investigation of Some Properties andDurability of Concrete in the Urumie Lake.
[13] K. Lau, A.A. Sagues, L. Yao, R.G. Powers, Corrosion performance ofconcrete cylinder piles, Corrosion 63 (4) (2007) 366–378.
[14] A.A. Sagues,Modeling the effects of corrosion on the lifetime ofextended reinforced concrete structures, Corrosion 59 (10) (2003) 854–866.
[15] Woosuk Ahn, Darrell Joque, and Amfinn Rusten , A Study on Corrosion
Resistance of Prestressed Marine Concrete Piles, corrosion 2003,page
nomber03239[16] Building research establishments news, Her Majesty's Stationanery
Office, London, Winter 1979[17] Haynes, H. H., "Permeability of Concrete in Sea Water" Performance ofConcrete in Marine Environment, ACI SP-65, edited by Malhotra, V. M.,
ACI, Detroit, MI, August 1980, pp. 22-38.
[18] C. Ftikos, G. Parissakis, A study on the effect of some ions contained inseawater on hydrated cement compounds, in: J.M. Scanlon (Ed.), Katharine
and Bryant Mather International Conference on Concrete Durability,
American Concrete Institute,1987, pp. 1651–1665, S P (100).
[19] Ahmad Shayan, Aimin Xu, Performance of glass powder as a pozzolanicmaterial in concrete: A field trial on concrete slabs, Cement and Concrete
Research 36 (2006) 457–468 December 2005[20] A. Shayan ,A. Xu, G. Chirgwin, H. Morris, Effects of seawater on AAR
expansion of concrete, Cement and Concrete Research, 40 (2010) 563–568[21] J.G. Wiebenga, Durability of concrete structures along the North Sea
coast of the Netherlands, American Concrete Institute SP65 (1980) 437–452.
[22] A.A. Sagues,Modeling the effects of corrosion on the lifetime ofextended reinforced concrete structures, Corrosion 59 (10) (2003) 854–866.
[23] Mehta,P.P. , Performance of Concrete in Marine Environment, ACI
SP-65,pp.1_20,1980397

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Fundamental Study on Particle Size Distribution of Coarse Materials by Image
Analysis
G.H.A.J.J. Kumara, K. Hayano and K. Ogiwara, Department of Civil Engineering, Yokohama National University,
Japan

ABSTRACT: Particle size distribution of materials is usually
evaluated by sieve analysis tests. In this research, an image
analysis technique using ImageJ is proposed to evaluate particle
size distribution of coarse material. On particular conditions, somedifference of particle gradation curves determined by sieve
analysis and image analysis were observed. Based on the results,
several aspects related to image analyzing are discussed in thispaper. They include appropriate evaluation of particle diameter inimage analysis, minimization of shadow effects appeared in image,effects of number of particles adopted for sieve analysis and image
analysis and so on.

Keywords: coarse materials, ImageJ, image analysis,
particle size distribution, sieve analysis
1. INTRODUCTION
Characterization of quality of materials is important toensure good use of resources from environmental andeconomic perspectives. Particle size distribution ofmaterials is one of the widely used tests in geotechnicalengineering to evaluate quality of materials. Sieve analysis
test has been the main method to determine particle size
distribution of coarse materials for many decades. However,recently, some researchers including Fernlund (1998) [1],Mora et al. (1998) [2], Kwan et al. (1999) [3], Banta et al.
(2003) [4], Fernlund (2005a) [5] and Fernlund (2005b) [6]
used some image analyzing techniques to evaluate particle
shape characteristics including particle size distribution.
In general, image analysis uses 2-D images. Banta et al.
(2003) [4] studied particle gradation curves by 2-D imagesof 4.75-25mm size limestone and found that image analysis
gave good results. However, mass of individual particles
were measured using a balance to compare gradation curves
by two methods. It was a time consuming process since themass of individual particles were measured. Kwan et al.
(1999) [3] used 2-D images of coarse aggregates to study
particle shape characteristics. However, rather thangradation curves, particle shape characteristics likeflakiness and elongation of aggregates were discussed.
Fernlund (1998) [1] studied particle form on sieve analysis
with 32-64mm size railroad aggregates using 2-D images.
Moreover, the difference of gradation curves determined bymass and no. of particles in hand sieve analysis tests werealso studied and found that gradation curve determined by
no. of particles underestimate that of by mass.
Mora et al. (1998) [2] compared gradation curves
determined by image analysis method and sieve analysis
test using 2-D images. However, rather than using area ofparticles measured directly from 2-D images, mass ofparticles were evaluated from 2-D images with someassumptions. The gradation curve overestimated gradation

curve determined by sieve analysis test. Therefore, a size
correction factor was assigned to get same gradation curves
as that of by sieve analysis test. Therefore, it is not clear
whether the assumptions in evaluating mass affected thedifference or image analysis had some effects on gradation
curves.
On the other hand, Fernlund (2005a) [5] measured all
three axes of particles of 10-50mm size granite while
manually changing positions of particles. 2-D images ofsame particles were taken twice to measure three axes.
Gradation curve determined by image analysis werecompared with that obtained by Danish Box. The results
showed that 2-D images when particles placed on the moststable location gave good results. However, it was a time
consuming process due to the fact that location of particles
had to be changed twice to obtain images to measure all
three axes.
As mentioned in literature, usually, gradation curves
were determined based on mass by image analysis
techniques than using direct measurement of area ofparticles measured from 2-D images. However, mass (orvolume) cannot be measured directly from 2-D images.
Moreover, it was seen that image analysis using direct
measurement of area of particles gave different results thansieve analysis. The difference of results would have been
due to several issues such as evaluation of particle diameter,
shadow effects appeared in image, effects of difference of
number of particles between sieve analysis and image
analysis and so on. However, they have not been fully
understood. Therefore, during this research, issues
mentioned here were examined using an image analysis
technique named ImageJ.
During this research, it was discussed how area ofparticles can be used to obtain gradation curves of coarse
materials varying from 2-19mm size as well as how no. ofparticles or random selection of particles from the main
sample affect gradation curve.
2. METHODOLOGY
Gravels ranging from 2-19mm were used for theexperiments. They are Andesite, originally produced in
Yamanashi prefecture in Japan.
2.1 Sieve analysis
Sieve analysis test was done according to JIS A 1204 [7].
JIS requirement for the materials used is 1.5kg. However, toobtain high accuracy for gradation curve, additional sieves
such as 6.7, 11.2 and 13.2mm were added in addition to 2,
4.75, 9.5 and 19mm sieves required by JIS method.
Moreover, sieve analysis test were conducted with handshaking.

399

2.2 Image capturing process
Particles were arranged manually such a way that particles
stand on their most stable positions. However, particles
were arranged without touching or overlapping to obtain
good quality images.
Particles were placed on transparent sheet during initial
analysis as shown in Fig. 1. Transparent sheets were used aslights were applied from both bottom and top of the sheet toeliminate shadow effects. The main light is back light. Top
light was applied to strengthen back light effect with areflective sheet placed as shown in Fig. 1. However, as it
was difficult to apply same light arrangement when the sizeof the sheet becomes large with higher no. of particles,
white or black color sheets were used on the table.

Transparent
sheet
Top light
Reflective
plate

Back light

Figure 1: Image capturing process
In Series A, whether there is any effect due to randomly
selection of particles were examined such a way thatparticles were placed on sheets varying from small size tolarge size as discussed in section 3.2 and given in Table 1.
Moreover, shadow effect was evaluated using black color
sheets in Series B as given in Table 2. Then, gradation
curves determined with white color sheets in Series A werecompared with that of by black color sheets in Series B.
Sieve analysis tests relevant to the sample used for image
analysis were done as given in Table 3.

2.3 Image analysis
All the images captured were evaluated in ImageJ software.
ImageJ can read many image formats including TIFF, GIF,
JPEG, BMP, DICOM and FITS. JPEG images were usedduring this research. It can calculate area and pixel value
statistics of user defined selections. Moreover, it canmeasure distance and angle as well. It supports standard
image processing functions such as contrast manipulation,
sharpening, smoothing, edge detection and median filtering.
More details on ImageJ can be found in Ferreira andRasband (2011) [8].
In ImageJ, it can convert original images into binary
images. It assumes that binary images have black objects
and white background. However, black background withwhite objects can also be obtained. Binary images are very
important to make some process like Erode, Dilate, Open,Close-, Fill Holes, Watershed and so on. Erode removes
pixels from the edge of black objects while Dilate adds
pixels to the edges of black objects. Open performs anerosion operation followed by dilation while Closeperforms a dilation operation followed by erosion. Fill
Holes fills holes in objects. Moreover, Watershed separates
or cuts touching particles. However, during this research,
particles were arranged such a way that they didn’t toucheach other. More details on different processes can be readon The Image User Guide given by Ferreira and Rasband(2011) [8].
Image analysis process conducted is shown in Fig. 2. As
shown in Fig. 2, firstly, pixel is converted into mm using ascale factor. Then, images captured were converted into
binary images. Figs. 3 - 7 show some of images and theirbinary images of samples with 100, 1100 and 3300 particles.
Depending on quality of the images judged by the operator,
no. of steps of Dilate, Fill Holes and Erode process were
decided. Moreover, no. of steps of Dilate and Erode werealways made equal to make there is no addition or removalof pixels from the real particles.

Pixel

Table 1: Image analysis – Series A
Case
No. of particles
Particle placed sheet
13300 x 1 image
Transparent
21100 x 3 images
White
3
100 x 33 images
White

Area

mm

Major
axis

Analyze
Pa rticles

Table 2: Image analysis – Series B
CaseNo. of particles
Particle placed sheet
11100 x 3 images
Black23300 x 1 image
Black

Crop
unnecessa ry
parts

Set scale

Ma ke
BinaryMinor
axis

Dilate

Fill
Holes

Erode

Figure 2: Image analysis process

Table 3: Details of sieve analysis tests
Case
No. of particles
a
100
b
3300
Images were captured by a Nikon D7000 camera. The
camera can measure up to 16 million pixels.

Since particles are in irregular shapes, during image
analysis, Bounding Rectangle and Fit Ellipse were used tofind areas enclosed by individual particles. Bounding
Rectangle gives the smallest area of a particle enclosed by arectangle while Fit Ellipse gives that of by an ellipse. Area,
major axis and minor axis were measured during image
analysis though many shape characteristics including
roundness, perimeter, area fraction and so on can bemeasured. More details on image processing techniques
available can be found in Ferreira and Rasband (2011) [8].

400Figure 5 (a): Image with 1100 particles on black sheet
Figure 3 (a): Image with 100 particles

Figure 5 (b): Binary image of Fig. 5 (a)Figure 3 (b): The binary image of Fig. 3 (a) image

Figure 6 (a): Image with 3300 particles on white sheet

Figure 6 (b): Binary image of Fig. 6 (a)Figure 4 (a): Image with 1100 particles on white sheet

Figure 7 (a): Image with 3300 particles on black sheet

Figure 4 (b): Binary image of Fig. 4 (a)Figure 7 (b): Binary image of Fig. 7 (a)401Sieve analysis tests use mass when evaluating cumulative
curve of particle size distribution whereas image analysis
can use either area of particles or no. of particles.
Some researchers including Banta et al. (2003) [4], Moraet al. (1998) [2] and Cheng (2000) [9] used mass
determined from volume of particles instead of area ofparticles. The assumptions in determining volume from 2-D
images would result in some errors. On the other hand, if
mass of particles are measured individually like in Banta etal. (2003) [4], that would be time consuming. Moreover,area of particles can easily be determined from 2-D images.
As a main objective of this research is to reduce time indetermining gradation curve, image analysis was done witharea of particles.
3.1 Effect of fitting shape on the evaluation of particle
diameter
As shown in Fig. 8, effect of shape of particles wasanalyzed using three different shapes such as sphere,
rectangle and ellipse. In this case, 100 no. of particles wererandomly selected by the operator and used. Diameter,
width and minor axis were used respectively for sphere,
rectangle and ellipse as the grain size, D. Though JGS
standard needs large amount of materials for sieve analysis(1.5kg which equals to 3300 particles for this material),
sieve analysis were conducted for the same samples of 100
particles used for image analysis for comparison purposes.
As it can be seen in Fig. 8, minor axis of ellipse gives theclosest results to that of by sieve analysis test among theshapes considered. It can also be seen that if rectangle orsphere shapes is used, gradation curves would overestimatethat of by sieve analysis tests. Therefore, hereafter, particles
were analyzed using minor axis of ellipse shape during the
research.
As particles are placed on the sheets with their moststable positions, it can be observed that 2-D images have
longest and intermediate axis while the shortest axis isperpendicular to the plane of images. Therefore, it can beconcluded that intermediate axis (minor axis on 2-D
images) should be examined when comparing the resultswith sieve analysis tests. Effects of three axes werediscussed in Fernlund (2005a) [5] and Fernlund (2005b) [6].
As discussed in Fernlund (2005a) [5] and Fernlund (2005b)[6], intermediate axis in 2-D images should give goodresults.

Percent passing (%)

100
SphereRectangle
EllipseSieve Analyse

80
60
40
20
0

2

3

4

5 6 7 8 9 10
Grain size, D (mm)203.2 Effects of random selection of particles
JGS standard requires minimum amount of materials
depending on the material’s maximum particle size. It was
found that JGS standard requirement is equivalent to 1.5kg
which equals to approximately 3300 particles for thegravels used in this research. However, when the operator
takes small sample (e.g. 100 particles) from the sample(3300 particles), it is not clear whether the operator gets 100
particles which would represent whole sample
appropriately. When there is large no. of particles available,
there might be some segregation and that would result inthe operator getting coarser particles at initial stages.
The effect of random selection from a sample of 3300
particles on the gradation curve was evaluated using many
samples of 100 particles and the results are shown in Fig. 9(a). In this case, 10 samples each with 100 particles wererandomly taken from the main sample of 3300 particles. As
shown in Fig. 9 (a), initially selected samples overestimate
gradation curve determined by sieve analysis test whilelater selected samples underestimate gradation curve
evaluated by sieve analysis test. However, it should be
noted that light arrangement and particle placed sheet weredifferent in these cases than in previous cases. Moreover, it
can be seen that gradation curve determined by sieve
analysis test stands at the middle of gradation curves
determined by image analyses.
100

Percent passing (%)

3. RESULTS AND DISCUSSIONSSA (3300 particles)
IA - Case 1
IA - Case 2
IA - Case 3
IA - Case 4
IA - Case 5
IA - Case 6
IA - Case 7
IA - Case 8
IA - Case 9
IA - Case 10

80
60
40
20
0

2

3

4 5 6 7 8 9 10
Grain size, D (mm)20Figure 9 (a): Effects of random selection of particles
Then, in the next analysis, the operator intentionally tried
to get a good sample to represent the main sample (3300
particles). All the particles (3300) were placed in 33
samples each with 100 particles and analyzed individually.
In these cases, compared to cases in Fig. 9 (a), the operator
purposely tried to select good samples of 100 particles thanjust randomly selecting like in Fig. 9 (a) since it wasobserved in Fig. 9 (a) that when particles are randomly
selected, the operator tends to get coarser particles initially.
The purpose was to evaluate whether the operator can selecta sample which would represent the main sample
appropriately. Even then, there is some difference between
the randomly selected samples and the main sample asshown in Figs. 9 (b) – (d).

Figure 8: Effect of fitting shape on the evaluation of particle
diameter402

Since taking many images as high as 33 to cover all theparticles would take some time, it was considered taking
less no. of images with all the particles. As given in Tables
1 and 4, to cover all the particles, it needs a very large size
sheet. However, shadow effects on particles closed to theboundary of the sheet could be a problem for very large
sheets.

100
IA - Case 1
IA - Case 2
IA - Case 3
IA - Case 4
IA - Case 5
IA - Case 6
IA - Case 7
IA - Case 8
IA - Case 9
IA - Case 10
IA - Case 11
SA - 3300 particles

Percent passing (%)

80
60
40

Table 4: Details of the size of the sheets
Case
Size of sheet
No. of images
1
20 x 13.2 cm
33
277 x 51.3 cm
33
130 x 86 cm
1

20
0

2

3

4

5 6 7 8 9 10
Grain size, D (mm)20

Fig. 10 shows gradation curves related to Case 2 where
three images with 1100 particles each were analyzed. As it
can be seen in Fig. 9 and Fig. 10, scattering of data becomeless when no. of particles in images are increased.

Figure 9 (b): Effects of random selection of particles –Cases 1- 11
100
IA - Case 12
IA - Case 13
IA - Case 14
IA - Case 15
IA - Case 16
IA - Case 17
IA - Case 18
IA - Case 19
IA - Case 20
IA - Case 21
IA - Case 22
SA - 3300 particles

60
40

100

20
0

2

3

4

5 6 7 8 9 10
Grain size, D (mm)IA - case 1
IA - case 2
IA - case 3
SA - 3300 particles

80
Percent passing (%)

Percent passing (%)

80

60
40
20

20

0

Figure 9 (c): Effects of random selection of particles –Cases 12- 22

23

4 5 6 7 8 9 10
Grain size, D(mm)20Figure 10: Gradation curves with 1100 particles

100
IA - Case 23
IA - Case 24
IA - Case 25
IA - Case 26
IA - Case 27
IA - Case 28
IA - Case 29
IA - Case 30
IA - Case 31
IA - Case 32
IA - Case 33
SA - 3300 particles

Percent passing (%)

80
60
40
20
0

2

3

4

5 6 7 8 9 10
Grain size, D (mm)20Figure 9 (d): Effects of random selection of particles –Cases 23- 33
As seen in Figs. 9 (b) – (d), it can be concluded that evenwhen the operator intentionally tries to take particles torepresent the main sample, there is still some difference
among particle sizes in each small samples. However, it canbe seen that gradation curve determined by sieve analysis
test stands at the middle of gradation curves determined byimage analyses. Therefore, it could be argued that if all theparticles used for sieve analysis test is considered for image
analysis, results could be good.

It can be seen in Fig. 11 that images with less no. ofparticles gives better gradation curves. As shown in Fig. 11,
in all cases (Series A), gradation curves by image analysis
overestimated gradation curve determined by sieve analysis
test. It could be assumed that the overestimation of
gradation curve determined by samples on large sheets
could be due to shadow effects. Moreover, as shown in Fig.
11, it can be observed that 33 images covering all theparticles give a gradation curve very similar to that of bysieve analysis test.
Error between D50 of gradation curves determined byimage analysis and sieve analysis test, e, was determined asgiven in (1). It was found the errors are 1.1, 7.3 and 8.8%
for Cases 1, 2 and 3 respectively. The error was always
positive, i.e. image analysis always overestimates thegradation curve determined by sieve analysis. Therefore, it
could be concluded that when very high accuracy neededon gradation curves, particles should be placed on a small
sheet such that shadow effects, probably at the edges of the
sheets, won’t be significant.
eD

50 , IA D50 , SA 

D50 , SAx 100 (%)

(1)

where D50,IA is D50 of gradation curve determined by image
analysis while D50,SA is that of by sieve analysis test.

403

4. CONCLUSION
100
IA - 100 particles x 33 images
IA - 1100 particles x 3 images
IA - 3300 particles x 1 image
SA - 3300 particles

Percent passing (%)

80
60
40
20
0
2

3

4 5 6 7 8 9 10
Grain size, D (mm)20

Fig. 11: Gradation curves with different no. of particles
3.3 Effects of color of particle placed sheet
It was also examined whether there is any effect of color ofparticle placed sheets on the gradation curves by lighteffects in Series B as given in Table 2.
Fig. 12 shows results of gradation curves determinedwhen particles were placed on black color sheets. It can beobserved that trend of gradation curves in Fig. 12 is similar
to that of Fig. 11 in which gradation curves by image
analysis overestimate that of by sieve analysis test.
Moreover, images with less no. of particles give better
gradation curves. However, in Fig. 12, gradation curve
determined by images with 100 particles are also included
for comparison purposes though that was determined withtransparent sheet as given in Table 1.
Moreover, it was determined the error, e, are 3.2 and
5.0% for Cases 1 and 2 respectively. Table 5 describes
comparison of errors with white and black color sheets. It
can be observed that black color sheets give better resultscompared to white color sheets, perhaps black color sheets
have less effect from lights than white color sheets.

Percent passing (%)

100
IA - 100 Particles x 33 images
IA - 1100 Particles x 3 images
IA - 3300 Particles x 1 image
SA - 3300 Particles

80

5. REFERENCES

60
40
20
0

Particle size distribution curve of coarse materials(2-19mm) were evaluated using image analysis, ImageJ.
2-D images of particles were used in image analysis.
Gradation curves by image analysis were compared withthat of by sieve analysis test. Finally, the following
conclusions were made,Since area of particles gives accurate gradation curves asthat of by sieve analysis test, image analysis technique used
was simple and less time consuming than sieve analysis test.
Moreover, it was also simple as mass from volume ofparticles are no need to be evaluated.
As particles were in irregular shapes, ellipse shape wasbetter to represent them properly than spheres or rectangle
shapes. When the shape of particles was represented bysphere or rectangle, gradation curves overestimate that ofby sieve analysis test.
Shadow effects can be eliminated with proper lightarrangements. That’s to say, if two lights are used
symmetrically, that would eliminate shadow effects verymuch.
All the particles should be considered to obtain accurate
gradation curves as taking a random sample from a large
sample of particles would not necessary represent the wholesample.
The errors of D50 were 7.3 and 8.8% with white color
sheets while they were 3.2 and 5.0% with black color sheets
for cases of 1100 and 3300 particles respectively. Therefore,
it can be concluded that black color sheets were better thanwhite color sheets to place particles. That’s to say, shadow
effects on black color sheets should be less than that ofwhite color sheet.
Error of D50 between two methods was 1.1% for images
with 100 particles on a transparent sheet. Moreover, they
were 3.2 and 5.0% for particles 1100 and 3300 respectively
when black color sheets were used. Therefore, when very
high accuracy on gradation curve is needed, small sheet
should be used to place particles to eliminate shadow
effects.

2

3

4 5 6 7 8 9 10
Grain size, D (mm)20Figure 12: Effects of color of particle placed sheet
Table 5: Details of error of D50 between SA and IA
Error, e (%)
No. of images
White sheet
Black sheet
3
7.3
3.2
1
8.8
5.0

[1] Fernlund JMR, The effect of particle form on sieve analysis: a test by
image analysis, Engineering Geology, Vol. 50, 1998, pp. 111-124.
[2] Mora CF, Kwan AKH and Chan HC, Particle size distribution analysis
of coarse aggregate using digital image processing, Cement andConcrete Research, Vol. 28, No. 6, 1998, pp. 921-932.
[3] Kwan AKH, Mora CF and Chan HC, Particle shape analysis of coarse
aggregate using digital image processing, Cement and Concrete
Research, Vol. 29, 1999, pp. 1403-1410.
[4] Banta L, Cheng K and Zaniewski J, Estimation of limestone particle
mass from 2D images, Power Technology, Vol. 132, 2003, pp.
184-189.
[5] Fernlund JMR, Image analysis method for determining 3-D size
distribution of coarse aggregates, Bull Eng Geol Environ, Vol. 64,
2005a, pp. 159-166.
[6] Fernlund JMR, Image analysis method for determing 3-D shape ofcoarse aggregate, Cement and Concrete Research, Vol. 35, 2005b, pp.
1629-1637.
[7] JIS A 1204, Test method for particle size distribution of soils, 2009, pp.
115-123 (in Japanese).
[8] Ferreira T and Rasband W, The ImageJ user guide 1.44, 2011.
[9] Cheng K, Optical Gradation for Crushed limestone aggregates, PhD
Dissertation, West Virginia University, 2000.

404

Environment

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

CALIBRATION AND SIMULATION OF RUNOFF MODELS FOR
DEVELOPING WATERSHED RESPOND KNOWLEDGE AT TROPICAL
AREA
Arien Heryansyah, Zulkifli Yusop. Faculty of Civil Engineering, Universiti Teknologi Malaysia



ABSTRACT: Knowledge of environmental respond at watershed
level is crucially important for good, scientifically and economically
accepted strategic planning design. Since most of the South East
Asian countries are tropical developing areas with short and few
historical data are available but rapidly change of land utilization,modeling approach is the only way to establish the knowledge ofwatershed responds to system change. Because, rainfall runoff
modeling can be use for flood mitigation, for water supply design,for controlling non-point source pollution, or for evaluating thewatershed with reasonable data requirement. However, the models
need specific local calibration therefore they valid on specificlocations and condition only. In this study a watershed tank model atseveral locations will be calibrated simultaneously then evaluatedaccording to their applicability, simplicity and possibility to become
regional tropical watershed runoff model. Simple regional scale ofwatershed model and better understanding of watershed respond toland use change for strategic planning were expected as the result of
the study.

Keywords: Tank model, land use change, tropical region,
calibration.
1. INTRODUCTION
Development and sustainability are sometimes usuallyconsidered as two words that contradict. Development is tomeets the needs of the present, while sustainability is topreserve the ability of future generations to meet their own
needs. However by using a lot of experience, data and local
wisdom, development and sustainability can be synergized.
But at developing country, hydrology and environment data
are always insufficient, or for short period only. A water
related development can have a significant impact on health,quality of life, economic development, natural resource, andthreatening the long-term viability of communities. On theother word, it is too many risks to conduct development andimpossible to synergize development and sustainability atdeveloping country without establish new way to interpret
short period of data.
Land uses change rapidly due to development, population or
economic pressure. As a hydrological unit, land uses has
important role in water related development. Interpretation offlood and drought of short period data without proper
understanding of the land uses pattern will cause designmalfunction: under or over estimate. As long as land use
pattern is considered in the interpreting the data and design,
the short period of data will not a barrier anymore. In this case,modeling approach is proposed to become tools for designer

or decision maker as well as new hydrological andenvironmental data observation station. Modeling can be used
for learning, and observe the impact of changing condition,
especially land use.
2. LITERATURE REVIEW
Hydrological runoff model have been developed from a needto analyze and solve specific hydrological problems, such as
to obtain useful outcome of the modeling exercise, to evaluatevariation of state-variables over space and time, and tounderstand internal flow processes. According to suitable
solution for those problems, hydrological model wasdifferentiated into distributed and lumped model. Normally,
distributed modeling associated with physically based
parameter and they are necessary if variation ofstate-variables over space and time are more important rather
than direct relationship between rainfall and discharge. On
the other hand, lumped model is more appropriate if the focus
is good agreements between calculated and discharge without
neglectbasichydrologicalprocesses,
suchasevapotranspiration (ET), infiltration, and water movements inthe soil. Calibrated parameters are used more often, as aconsequence of simplification.
The lumped model is suitable in the typical application of ahydrological model on simulation of discharges under
stationary catchments conditions [1]. The most famous of the
lumped model is Tank Model. It was broadly use forsimulated rainfall-discharge relation in various watershed
condition [2]. Tank model also can be used as specific tools,
such as for evaluating the retention capacity of watershed.
Another researcher use diffusive tank model with 10 minutes
interval for flood analysis [3] or for run off analyses in paddy
field [4]. Watershed modeling that thank model as basis has
succeeded to give right figure to decision maker at Cidanau
Watershed [5].
Areal Non-points Source Watershed Environment Response
Simulation model, ANSWERS, is one of famous distributed(grid based) hydrological model. Its primary application wasenvisioned to be planning and evaluating various strategies
for controlling nonpoint source pollution from intensively
cropped areas [6]. Similarly, [7] proposes Hydrological
Similar Unit (HSU), a kind of distributed type modeling, totake into consideration land use contribution on hydrologicalprocesses. And recently, [8] succeeds to identify hydrologicalprocesses at landscape zone at sub tropic area. All these threemodels employ land use role in hydrological processes but

347

these models were considered to be complex and required a
lot of data.
However, all the models valid on specific locations only andthey need calibration for each location. Model applications at
tropical area are should be more simple and required lesscomplicated data. Therefore in this study, the samehydrologic model on several watersheds will be evaluatedaccording to their applicability, simplicity, and reasonability
relationship between model parameter and physical based
parameter. On top of that, a kind of regional model fortropical area was expected to be established.

Table 1. The land use scenario of each watershed
Condition

Existing

SLU-1

2 STUDY AREA
The research was conducted at two watershed at Java Island(Cidanau and Citarum watershed) and two watershed at
Sumatra Island (Krueng Aceh and Musi Watershed). Cidanau,
Citarum, and Musi are located on South hemisphere, while
Krueng Aceh Watershed is located on Northern Hemisphereas presented on Fig.1. They are considered to be representing
tropical region.

SLU-2

SLU-3

a

b

SLU-4

c d

Fig 1. Study area: (a) Krueng Aceh; (b) Musi; (c) Cidanau; and (d) Upper
Citarum Watershed
SLU-5

3 METODOLOGYCollect the rainfall, discharge and land use data from several
Watershed / Sub watershed are the first step of this research.
Then continued with water balance analysis, model
development, calibration, sensitivity analysis and simulation.
Water balance was conducted to understand the hydrologicalcycle (1) and to verify the hydrologic data.
Rainfall = Discharge+ET+Storage

SLU-6

(1)
SLU-7

Four layer lumped tank model was employed. This lumped
tank model modeling replicated five typical hydrologicalprocesses at tropical area, i.e. evapotranspiration, infiltration,
surface flow, inter flow, and base water flow. Each process
was represented by calibrated parameter in calibration step
such as CR for coefficient of runoff (surface, inter, and baseflow), and CP for coefficient of percolation.
Calibration was conducted using automatic trial and error. It
generates random value for CR and CP of each land use type,
and then calculates values for the watershed proportionallywith land use. The optimum values were evaluated using
Model Efficiency (ME) and Mean Relative Error (MRE). Bias
Error (BE) was calculated also to express the balance
condition between observed and calculated discharge.

SLU-8

Watershed

1

2

3

4

5

6

7

Cidanghiang

21.0%

25.9%

0.0%

8.3%

41.8%

3.0%

0.0%

Mandalawangi

6.4%

12.9%

0.0%

56.2%

24.4%

0.1%

0.0%

Peusar

7.5%

34.4%

9.0%

36.1%

11.0%

1.8%

0.2%

Cidanau

6.8%

31.5%

8.2%

41.1%

10.2%

2.0%

0.3%

KruengAceh

56.9%

0.2%

0.0%

0.3%

37.6%

4.7%

0.2%

Keleker (Musi)

3.7%

0.0%

1.4%

85.7%

1.8%

7.3%

0.1%

Upper Citarum

15.5%

20.6%

0.0%

5.4%

48.1%

10.3%

0.1%

Cidanghiang

33.9%

25.9%

0.0%

1.9%

35.3%

3.0%

0.0%

Mandalawangi

22.6%

12.9%

0.0%

48.0%

16.3%

0.1%

0.0%

Peusar

10.5%

34.4%

9.0%

34.6%

9.5%

1.8%

0.2%

Cidanau

9.6%

31.5%

8.2%

39.7%

8.8%

2.0%

0.3%

KruengAceh

87.5%

0.2%

0.0%

0.0%

7.3%

4.7%

0.2%

Keleker (Musi)

10.7%

0.0%

1.4%

80.4%

0.0%

7.3%

0.1%

Upper Citarum

64.3%

20.6%

0.0%

0.0%

4.7%

10.3%

0.1%

Cidanghiang

1.7%

25.9%

0.0%

14.8%

48.2%

9.4%

0.0%

Mandalawangi

1.5%

12.9%

0.0%

57.8%

26.0%

1.8%

0.0%

Peusar

5.4%

34.4%

9.0%

36.8%

11.7%

2.5%

0.2%

Cidanau

4.9%

31.5%

8.2%

41.7%

10.8%

2.6%

0.3%

KruengAceh

0.2%

0.2%

0.0%

53.3%

37.6%

8.5%

0.2%

Keleker (Musi)

0.0%

0.0%

1.4%

88.9%

1.8%

7.8%

0.1%

Upper Citarum

1.0%

20.6%

0.0%

15.1%

48.1%

15.1%

0.1%

Cidanghiang

0.0%

25.9%

0.0%

15.3%

48.8%

10.0%

0.0%

Mandalawangi

0.0%

12.9%

0.0%

58.3%

26.5%

2.3%

0.0%

Peusar

5.1%

34.4%

9.0%

36.9%

11.8%

2.6%

0.2%

Cidanau

4.7%

31.5%

8.2%

41.8%

10.9%

2.7%

0.3%

KruengAceh

3.9%

0.1%

0.0%

0.3%

90.5%

4.9%

0.2%

Keleker (Musi)

0.0%

0.0%

1.4%

78.8%

12.3%

7.3%

0.1%

Upper Citarum

0.5%

6.1%

0.0%

5.4%

57.8%

30.0%

0.1%
0.0%

Cidanghiang

0.0%

0.0%

0.0%

15.3%

61.7%

22.9%

Mandalawangi

0.0%

0.0%

0.0%

58.3%

33.0%

8.7%

0.0%

Peusar

5.1%

30.8%

9.0%

36.9%

13.6%

4.4%

0.2%

Cidanau

4.7%

28.2%

8.2%

41.8%

12.5%

4.3%

0.3%

KruengAceh

11.5%

0.0%

0.0%

23.0%

60.3%

4.9%

0.2%

Keleker (Musi)

0.0%

0.0%

1.4%

64.8%

22.7%

11.0%

0.1%

Upper Citarum

0.0%

10.9%

0.0%

19.9%

48.1%

20.4%

0.6%
0.0%

Cidanghiang

0.0%

0.0%

0.0%

0.0%

61.7%

38.3%

Mandalawangi

0.0%

0.0%

0.0%

0.0%

33.0%

67.0%

0.0%

Peusar

5.1%

30.8%

9.0%

28.6%

13.6%

12.7%

0.2%

Cidanau

4.7%

28.2%

8.2%

34.2%

12.5%

11.9%

0.3%

KruengAceh

53.1%

4.0%

0.0%

26.8%

3.5%

12.3%

0.2%

Keleker (Musi)

0.5%

0.0%

1.4%

54.3%

1.8%

41.9%

0.1%

Upper Citarum

5.8%

1.3%

0.0%

5.4%

53.0%

34.4%

0.1%

Cidanghiang

0.0%

0.0%

0.0%

0.0%

0.0%

100.0%

0.0%

Mandalawangi

0.0%

0.0%

0.0%

0.0%

0.0%

100.0%

0.0%

Peusar

5.1%

30.8%

9.0%

23.6%

7.3%

24.0%

0.2%

Cidanau

4.7%

28.2%

8.2%

29.7%

6.8%

22.2%

0.3%

KruengAceh

56.9%

15.4%

0.0%

3.4%

14.9%

7.7%

1.7%

Keleker (Musi)

0.5%

0.0%

0.4%

43.8%

1.5%

53.7%

0.1%

Upper Citarum

0.5%

20.6%

0.0%

10.3%

28.9%

38.7%

1.1%
0.0%

Cidanghiang

29.4%

0.0%

0.0%

0.0%

67.6%

3.0%

Mandalawangi

62.6%

0.0%

0.0%

0.0%

37.3%

0.1%

0.0%

Peusar

30.9%

34.4%

9.0%

12.7%

11.0%

1.8%

0.2%

Cidanau

28.2%

31.5%

8.2%

19.7%

10.2%

2.0%

0.3%

KruengAceh

60.7%

0.0%

0.0%

0.3%

37.8%

0.9%

0.2%

Keleker (Musi)

3.8%

0.0%

1.5%

36.5%

50.7%

7.6%

0.0%

Upper Citarum

15.5%

10.9%

0.0%

0.6%

48.1%

24.7%

0.1%

Cidanghiang

93.8%

0.0%

0.0%

0.0%

3.2%

3.0%

0.0%

Mandalawangi

99.2%

0.0%

0.0%

0.0%

0.7%

0.1%

0.0%
0.2%

Peusar

40.4%

34.4%

9.0%

12.7%

1.6%

1.8%

Cidanau

36.8%

31.5%

8.2%

19.7%

1.6%

2.0%

0.3%

KruengAceh

56.9%

4.0%

0.0%

16.2%

22.5%

0.2%

0.2%

Keleker (Musi)

25.7%

0.0%

0.4%

17.5%

50.0%

6.3%

0.1%

Upper Citarum

29.9%

20.6%

0.0%

0.6%

38.5%

10.4%

0.0%

Note: (1) Forest, (2) Paddy Field, (3), Swamp Forest, (4) Plantation, (5)Upland Area, (6) Settlements, and (7) Open water

Sensitivity analysis was used to identify the most influence
land use and/or parameter, while simulation used forunderstanding the land use management. Seven types of land
uses such as forest, swamp forest, paddy field, plantation,348

upland area, and settlements, and open water were considered
suitable to represent most frequent land use at tropical area.
Several reasonable scenarios were developed to learn how tomanage the watershed using land use change. They are basedon the change of forest into plantation, upland, or settlement
as worst scenario; compensate housing area using lake (openwater) as moderate scenario; maintain forest and upland asconservative scenario. The land use change scenarios arepresented on Table 1.

watershed, detail information about cropping pattern isrequired to replicate the water loss in form of ET during ayear. However, since there are at least 3 paddy field dominantwatersheds with sufficiently good performance such as
Cidanau and Krueng Aceh, the model was considered to besuitable for land use change simulation.
Table 4. The model performance(Sub) Watershed
Cidanghiang
Mandalawangi
PeusarCidanau
KruengAceh
Keleker (Musi)
Upper Citarum

4 RESULTAND DISCUSSION
The water balance analysis shows that the watersheds havealmost similar water balance pattern, and reasonable water
movement (Table 2). The watersheds with negative storage
mean water shortage occur during dry season. Most of themlocated at near the coastal area, such as Peusar, Cidanau, andKrueng Aceh Watershed. On the other hand, several spring
locations were found at the downstream of watershed withpositive storage. It means the storage, in this casegroundwater, was released at downstream area.
Table 2. The water balance analysis(Sub) Watershed

RainfallCidanghiang
Mandalawangi
PeusarCidanau
Krueng Aceh
Keleker (Musi)
Upper Citarum

Discharge

ET

Storage

14%
35%
44%
48%
52%
68%
68%

51%
42%
66%
59%
72%
12%
59%

35%
23%
-10%
-6%
-24%
21%
-27%

2607.03664.76531.8
19937.5822.0
2706.93001.2The optimum calibrated parameters for each land use were
evaluated using several assumptions, such as general
understanding of hydrological processes, and physicalcharacteristic of the land uses. Runoff coefficient first layer
should be higher than second layer, as well as percolation
coefficient at first and second layer. Then runoff coefficientof forest should be lower than paddy field, plantation, or
upland area. And also, runoff settlement area and open water
should be higher than plantation, or upland area. Optimum
and evaluated parameter are presented on Table 3.

1

2

3

4

5

6

7

CRO
CR1
CR2
CR3
CP1
CP2

7.13E-04
9.75E-06
9.16E-06
9.07E-06
1.68E-02
1.37E-02

1.17E-03
1.91E-05
6.18E-06
6.12E-06
6.45E-03
3.63E-05

2.25E-03
1.72E-04
6.12E-06
6.06E-06
3.61E-06
1.19E-06

2.44E-03
4.43E-04
6.07E-06
6.01E-06
1.06E-06
4.77E-07

3.95E-02
4.43E-04
2.82E-06
9.16E-07
7.78E-07
5.29E-08

4.36E-02
6.53E-03
2.09E-06
3.86E-07
1.07E-12
8.27E-13

6.10E-01
6.04E-01
4.77E-07
1.65E-07
3.71E-13
1.40E-13

Note: (1) Forest, (2) Paddy Field, (3), Swamp Forest, (4) Plantation, (5)Upland Area, (6) Settlements, and (7) Open water

The model performance for almost all the watersheds are
sufficient, as presented on Table 4. Only one watershed(Upper Citarum) has very low performance. Probably it
caused by improper interpretation of land use classification.
At upper Citarum watershed, paddy field is dominant land use
since it occupies more than 20% of total area in addition to
39% of upland paddy field. For paddy field dominant

MRE

BE

0.43710.25880.37190.52800.35350.67000.4637-0.1340
0.10660.03580.03290.1869-0.10640.5232

Sensitivity analysis result is presented on Table 5. Forest,
paddy field, plantation and upland are considered as mostinfluencing land use, while runoff coefficient at 1st and 3rdlayers are consider as most influencing parameter. It meanswatershed manager should be thinks deeply before change thefraction of the most influencing land use. On the other words,
the most influencing land uses are required to be maintained
or to be increased to prevent water resources problem. This is
in line with the opinion of naturalists and conservative.
However in term of engineering view of point there arepossibilities to manage the land use using structure also. To
convert forest into others is acceptable as long as keep therunoff coefficient, especially at 1st and 3rd layers, at the samelevel with forest.
Table 5. The result of sensitivity analysis

Tabel 3. The Optimum and evaluated parameter
Parameters

ME
0.44320.45250.56600.67200.58910.56670.1845

Parameter

Land Use

Runof Coefficient-1
Percolation Coefficient-1
Runof Coefficient-3
Runof Coefficient-0
Runof Coefficient-1
Runof Coefficient-1
Percolation Coefficient-1
Runof Coefficient-3
Runof Coefficient-2
Runof Coefficient-1
Runof Coefficient-3
Runof Coefficient-3
Runof Coefficient-0
Runof Coefficient-1
Runof Coefficient-0
Runof Coefficient-3
Runof Coefficient-1

ForestForestForestOpen Water
Open Water
Paddy Field
Paddy Field
Paddy Field
Plantation
Plantation
Plantation
Settlements
Settlements
Swamp Forest
UplandUplandUplandChanging
Frequency
4
16
18
6
15
8
12
17
3
11
11
5
6
4
14
14
16

Simulation result as presented on Table 6 and Table 7 showthat Forest has important role to reduce the maximum
discharge and increase the minimum discharge. Contrary,
settlement has inverse role of forest. They show also that insome case (Cidanau), convert forest into plantation also canmaintain the maximum discharge at existing level. On top of
that, better understanding of watershed management in termof land use change, especially minimum and maximum
discharge, were expected to be achieved.

349

Tabel 6. The impact of land use change on minimum discharge.
Minimum
3

Discharge (m /s)
Cidanghiang
Mandalawangi
PeusarCidanau
Krueng Aceh
Keleker (Musi)
Upper Citarum[7]

Existing SLU-1 SLU-2 SLU-3 SLU-4 SLU-5 SLU-6 SLU-7 SLU-8

[8]
0.1
1.0
0.6
0.2
1.4
0.0
0.3

0.1
0.9
0.6
0.2
2.2
0.0
0.8

0.2
0.8
0.5
0.2
0.2
0.0
0.2

0.2
0.9
0.5
0.2
0.1
0.0
0.1

0.0
0.1
0.4
0.2
0.5
0.0
0.0

0.0
0.0
0.4
0.1
1.7
0.0
0.1

0.0
0.0
0.4
0.1
1.1
0.0
0.1

0.1
1.2
0.8
0.4
1.8
0.0
0.3

0.1
1.6
0.9
0.6
1.8
0.0
0.5

Tabel 7. The impact of land use change on maximum discharge.
Maximum
3

Discharge (m /s)
Cidanghiang
Mandalawangi
PeusarCidanau
Krueng Aceh
Keleker (Musi)
Upper Citarum

Existing SLU-1 SLU-2 SLU-3 SLU-4 SLU-5 SLU-6 SLU-7 SLU-8
6.9
16.6
174.7
576.5
680.8
43.0
5.5

4.3
8.1
19.5
26.1
31.1
45.0
7.9
14.5
30.0
35.0
56.8
69.0
166.3 194.9 197.9 205.5 221.7 243.1
915.5 1086.9 1098.1 1139.7 1165.2 1197.4
252.3 1242.0 1070.5 765.8 369.4 1519.6
31.9
89.7
90.6
93.0
93.7
97.8
4.0
7.8
9.8
16.2
9.1
24.0

5.7
6.0
130.7
456.7
318.7
32.7
6.8

3.4
4.4
121.5
355.3
320.3
22.2
3.2

5 ACKNOWLEDGMENTS
This research was a part of the Research University Programof Universiti Teknologi Malaysia, 2011. I wish to express mysincere gratitude to Dean of Water research Alliance for his
support in conducting this research. I also wish to thank to mycolleges Dr Yanuar and Prof Rizaldi at Bogor Agricultural
University for their continuous advises and other kinds ofassistance.
6 CONCLUSION
The results of this study are summarized as follows: Calibration and simulation of runoff model were donewith sufficient performance.
 This lumped tank model work properly on at least 2 sites
near to equator in this study area. Therefore this model
can be proposed as regional model on tropical region.
 Better watershed management was expected to beachieved since impact of land use on minimum andmaximum discharges are understood.
7

REFERENCES[1]

Refsgaard J.C and Knudsen J, Operational validation andintercomparison of different types of hydrological models, Water
Resources Research, 1996: 32 (7), 2189-2202
Sugawara, M. On the Analysis of the Runoff Structure of Several
Japanese Rivers. Japanese Journal of Geophysics, 1961: Vol. 4, No. 2.
Hayase, Y., Kadoya, M., Diffusive tank model for flood analysis and its
fundamental characteristics: runoff analysis by the diffusive tank model
in low-lying drainage basin. Trans. Jpn. Soc. Irrigat., Drainage andReclam. Eng., 1993: 165, 75–84 (in Japanese).
Chen, R.S., and Pi, L.C. Diffusive tank model application in
rainfall-runoff analysis of upland fields in Taiwan. Agricultural Water
Management, 2004: 70, 39–50.
Heryansyah Arien., Goto Akira., Mizutani Masakazu., and Yanuar
Muhammad JP. Development of Watershed Water Quality Model inIndonesia, Transactions of JSIDRE, 2007: No.251 (Vol.75-5), 97-105.
Bouraoui, F. and T. A. Dillaha. ANSWERS-2000: Runoff and sedimenttransport model. Journal of Environmental Engineering, ASCE, 1996:
122(6):493-502.

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T. Karvonena, H. Koivusaloa, M. Jauhiainena, J. Palkob, K. Wepplingc.
A hydrological model for predicting runoff from different land use
areas. Journal of Hydrology 217, 1999: 253–265.
Yang, Y., Xiao, H., Wei, Y., Zhao, L., Zou, S., Yin, Z., Yang, Q.,
Hydrologic processes in the different landscape zones of Mafengou
River basin in the alpine cold region during the melting period, Journalof Hydrology, 2011. Published Online.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Effect of Geological Succession on Macrophyte and Microbiota in Aquifer
Ecosystem in Urban Coastal Zone
Kazuhito MURAKAMI* and Akiko INOUE-KOHAMA***Dept. of Life and Environmental Sciences, Chiba Institute of Technology, 2-17-1 Tsudanuma, Narashino, Chiba
275-8588 Japan**Dept. of Environmental Information Engineering, Tohoku Institute of Technology, 35-1 Kasumicho, Yagiyama,
Sendai, Miyagi 982-8577 JapanABSTRACT: This study was conducted to investigate the effect ofgeological succession of sediment on macrophyto and microbiota inaquifer ecosystem. The research field was selected as Yatsu Tidal
Flat where located in urban coastal zone in Tokyo Inner Bay. Thistidal flat has been registered under Wildlife Protection Area of Japan
in 1988 and the Ramsar Convention in 1993, and just as the tidal flatthat was left from reclamation of waterfront zone of urban area. It isvery important to obtain some information about these ecosystemcharacteristics of such nature left in urban area, because creation of
artificial tidal flat is recognized as one of important enterprises toconstruct urban ecosystem. In recent years, a large chlorophyceae,
Ulva spp. (mainly Ulva pertusa and Ulva japonica) which isrecognized as a biotic indicator of eutrophicated sea area, has
become observed in Yatsu tidal flat, and it is considered that Ulva
spp. give a large effect to water quality. Ulva spp. has high absorbent
potential of nutrient salts at its growth phase, so this macrophyte isconsidered to place the microbiota such as zooplankton andphytoplankton under its control. In the first place, the geologicalsuccession as muddy to sandy occurred because of water qualitychange as fresh water to sea water. This water change resulted fromthe decrease of drainage with the sewer system service in basin area,and lead oligotrophication of the sediment. Ulva spp. is one of thebiological indicater for sea water and sandy sediment andeutrophicated environment. Because of the irregular growth of Ulva
spp., the quantity of phytoplankton decreased drastically, and thedominant species of phytoplankton and zooplankton transited from
ideal to unsuitable food organisms for macrobenthos which isimportant food for the migratory birds, and the production potentialof this aquifer ecosystem became in fragility.

Keywords: urban coastal zone, macrobenthos, macrophyte,geological succession, Yatsu tidal flat
1. INTRODUCTION
Tidal flat is located in the boundary zone between land andsea area. Therefore, the environmental factors are changing
periodically and very rapidly, according to supply of fresh
water from river and tide action. On the other hand, the algal
production speed and rate, such as diatoms and/or flagellums
on surface of sediment that is forming bio-film, is very high,because of much nutrient salts are supplied from both landand sea areas. Unique and precious ecosystem is constituted,
and birds and fishes gather to take feed at the same time. Tidal
flat is very important place as familiar nature, for not only
wild lives in tidal flat but also human in surrounding area.

And it is the place where for recreation, and to obtain marine
resources [1]. Although the area is decreased by reclamation
because of tidal flat is shallow sea and therefore it is easy toreclaim, the importance of wetland including tidal flat isindicated recently, and Ramsaor Convention whose purpose
is to protect migratory is adapted at the year 1971. Eleven
wetlands are registered for the convention in Japan [2].
Yatsu Tidal flat where is registered under Ramsaor
Convention, is reclaimed its circumference, and shut out from
Tokyo Inner Bay for over 20 years. And environment of thistidal flat is terrifically changed, for instance decrease of
inflow load and pure water originate in domestic wastewaterby improvement of public sewerage system. Furthermore, it
is recognized apprehensions in recent years that there ispossibility to occur the runoff of sediment by tidal current.
From analysis for the environmental succession and factors of
this unique tidal flat, some essential suggestions forconstruction artificial tidal flat in urban area is expected to beobtained.
In this paper, the study was conducted to investigate theeffect of geological succession of sediment on macrophyto
and microbiota in aquifer ecosystem.
2. OUTLINE OF YATSU TIDAL FLATIn the extreme end of the Tokyo Inner Bay, an
un-reclaimed Tideland of 40 ha is left which is now called as
Yatsu Tidal Flat (Yatsu Higata) as shown in Fig.1. Sea water
comes in and out with the ebb and flow of the bay. Large
numbers and different kinds of shorebird and migratory bird
come here each season. They are giving tremendous
relaxation to the community in the midst of the urban area.
Yatsu Tidal Flat was designated as an official site of Ramsar
Convention (Convention on Wetlands of International
Importance Especially as Waterfowl Habitat) in 1993.
In 1898, it was a shoal beach and was used for salt
manufacturing. In 1924, Keisei Rail Road (Keisei Dentetsu)
reclaimed the area and made a play land Yatsu Play Land
(Yatsu Yu-en). The remaining shore served for swimmingand clam fishing. The community had great fun out of it. In
1979, Tokyo Bay Reclamation project encroached the shore
and Yatsu Tidal Flat came to be in danger of landdevelopment. Groups and community people stood against

411

for its protection (Fig.2). In 1984, the conservation of the
Yatsu Tidal Flat was then included in the construction projectof local area public facility. Later in 1988, Yatsu Tidal Flat
was designated as the Natural Wildlife Protection Area. In
1982, Yatsu Play Land was closed due to the opening of theTokyo Disney Land in the nearby town. In 1993, Yatsu Tidal
Flat was designated as the Ramsar Site by The 5th
Conference of the Contracting Parties to the Convention on
Wetlands. In 1994, Observation Center was opened. In 1996,
Yatsu Tidal Flat joined to The East Asian and Australasian
Shore Bird Reserve Network in The 6th Conference in
Brisbane, Australia. In 1997, Narashino City enacted The
Day of Yatsu Tidal Flat for celebration of Yatsu Tidal Flat
conservation. In 1998, Narashino City and Brisbane City
Council agreed to The Affiliation Agreement of their
Wetland Conservation.
In Yatsu Tidal Flat, during the flow tide, the tidal flat is in
the water (0.8m depth in average), and during ebb tide, thetidal flat appears. The tidal flat is connected to Tokyo Inner
Bay through two canals, and the water flows in and out. So
many shorebirds gather and feed themselves, and so many
creatures are inhibiting such as crabs and invertebrates on
which the birds live. Throughout a year, many kinds of birds
are seen in Yatsu Tidal Flat. Seasonally there comes huge
number of seasonal migratory birds. For the shorebirds flying
between Siberia and Australia, Yatsu Tidal Flat is theirimportant intermediate rest place after and before their long
journey [3].

Fig.1 Location of Yatsu Tidal Flat

Fig. 2 Succession of environmental condition around Yatsu
Tidal Flat

3. RECENT PLOBLEMS IN YATSU TIDAL FLAT
Since 1983, Ulva sp., a marine macrophyte(chlorophyceae) which forms the green tide with its irregular
growth, has been observed and increased in its individual
number. The occupied zone in Yatsu Tidal Flat has expanded,that is, 6.8 ha in 1995, 13 ha in 1993, 20 ha in 2000, and 40 ha(completely occupied) after 2005 (Fig.3) [4]. The main
species of Ulva spp. was identified as U.pertusa andU.japonica. The reasons of this phenomenon, aneutrophication of water quality, increasing of seawater supply,and sediment succession from mud to sand have beenreported (Fig.4) [5].
Ulva spp. is recognized as one species of large green algaethat like sand or sandy-mud sediment condition rather than
mud sediment condition [6][7]. That is, it is thought thatgrowth of Ulva spp. is brought about by existence ofsufficient nutrient salts and progress of chlorinization ofwater and sandification of sediment in this tidal flat. Fromthese outcomes, the rise of a ratio (sand / mud) and that of (seawater / fresh water) can be considered as one of reasons forthe irregular growth of Ulva spp..
Sediment condition of the area where Ulva spp. is muchgrown (ORP ： -200 ～ -250mV) is presenting the black
reduction state compared with the sediment of the area where
Ulva spp. is not observed (ORP：-50～-100mV). So, in such
reduction area, there cannot be observed any macrobenthos
412

Fig.3 Succession of occupied zone by Ulva spp.

Fig.4 Ulva spp. multiplicated in Yatsu Tidal Flat

that is supporting a tidal flat ecosystem as main predator [8].
It is considered that Ulva spp. decayed and made anaerobic
Sediment state after its irregular growth. In comparison with
1984 and 1995, the total species number of macrobenthos has
increased from 13 to 39, however, the total individual number
has clearly decreased, and no living point was also observed[9]. That is, it is suggested that biological production
mechanism has been changing from “large quantity of small
species” condition to “small quantity of large species”
condition, namely, change to brittle bio- production state.
Thus, a change of phytoplanktonic flora which plays animportant role on food chain of the tidal flat ecosystem as theprimary producer, and that of zooplanktonic fauna as theprimary consumer leads a change of macrobenthic fauna as
the food source of the migratory birds. The decrease of
migratory birds attracts very few customers who visit to
Yatsu Tidal Flat to relax, and this leads decay of recreational
function, that is, the ecosystem service.
4. GEOLOGICAL SUCCESSION IN YATSU TIDALFLAT
In Yatsu Tidal Flat, the mud area has decreased from 20 ha
to 15 ha, sandy-mud area has increased from 10 ha to 18 ha,
and Ignition Loss (I.L.) of sediment, that indicates thequantity of organic matters in sediment, has also been
decreased (Fig.5). Moreover, although the thickness of the
mud layer in Yatsu Tidal Flat was 25-200cm in 1984, it isdecreased rapidly 0-50cm in 1995. This means sedimentcondition of Yatsu Tidal Flat is changing from mud to sand,and it was suggested at the same time that flow out of
sediment, especially mud layer from this tidal flat to Tokyo
Inner Bay.
The amount of inflow load that is flown from its basin area
to Yatsu Tidal Flat is decreased 16.8% for these 12 years from1983 to 1995 (Fig.6), and the diffusion of the seweragesystem in basin area is 85.3% in 2010. The decrease isbecause of the amount of drainage inflow from the household
decreased with improvement of public sewerage system of abasin area. Furthermore, improvement of public sewerage is
not only reduction of inflow loads, but also brought about theresult of reduction of the amount of fresh water supply.
Chloride ion concentration in Yatsu-Funadamari located inmost inner part of Yatsu Tidal Flat continues rising (Fig.7).
This phenomena namely means the water quality in Yatsu
Tidal Flat is changing from fresh water to sea water.
At Yatsu Tidal Flat, the inflow loads has decreased as
improvement of public sewerage system that is started since1990. As the result, water quality has changed from fresh
water to sea water and sediment condition has also changed
from mud to sand, and it is caused the irregular growth of
Ulva spp., which is the indicator of eutrophicated state. Thatis, as shown in figure 6, at the first step of succession, a tidal
flat just like tideland lake is formed artificially withreclamation and development of its circumference, and thewater quality became fresh water and sediment condition
became muddy, in gradually. So it is formed ecosystem thatFig.5 Comparison of sediment quality in 10 years

Fig.6 Change of popularization of sewerage and amount ofloads

Fig.7 Change of chloride ion concentration

has high-bio-production potential. However, in recent years,
supply of fresh water and organic matter from land area has
decreased with improvement of public sewerage system, so
chloronization of water quality and sandification of sedimentcondition have progressed, and bio-production potential has
reduced. Moreover, tide action becomes more influence, suchas flow out of sand from Yatsu Tidal Flat. From theseoutcomes, it was made clear that Yatsu Tidal Flat has beenchanging to the foreshore tidal flat again with remaining the
topographical characteristics as tideland lake.
5. SUCCESSION OF MICROBIOTA
Regarding to phytoplanktonic flora which plays animportant role on the food chain as the primary producer of
the tidal flat ecosystem, microscopic observation has been
conducted for several years (Fig.8). As results obtained fromthe observation, the following outcomes were obtained. 1) As

413

quality of phytoplanktonic flora, seasonal change of species
composition of phytoplankton is not so large and its species
number was 50-60 species through a year, 2) As quality ofphytoplanktonic flora, individual numbers of phytoplankton
was larger in summer and smaller in winter, and supplied
from Tokyo Inner Bay to Yatsu Tidal Flat throughout twochannels, called Yatsu river in the eastern side and Takase
river in the western side, with the ebb and flow of the bay, 3)
The diversity index, calculated by Shannon Index, ofphytoplankton is larger in winter and smaller in summer, andthe possibility of simplisation of phytoplanktonic flora wasoccurred by outflow from Yatsu Tidal Flat to Tokyo Inner
Bay, 4) Unique ecosystem has been constructed in thisenclosed tidal flat because of its closure circumstance which
greatly influencing to phytoplanktonic flora.
The macrobenthos mainly observed in Yatsu Tidal Flat
were Cinidaria, Mollusca, Annelida, Arthropoda and
Lophophorata. As biotic indicator of organic polluted
condition, Annelida was much observed in mud sedimentzone. Macrobenthic fauna is considered poor in un-polluted
environment and rich in polluted environment in general [10] .
But in further polluted environment, the species andindividual number of macrobenthos become to decrease, andmacrobenthic fauna become poor again as a result. Thistendency is not only in case of macrobenthos but also in mostof biotic indicator, for example, phytoplankton, water glass,and so on. The wealthy nature means a state that co-exist withmany species, many individuals, and without one or somedominant species. Yatsu Tidal Flat was considered to be in
the state which environmental condition is shifting from thefurther polluted condition to the polluted condition with theprogress of sandy sediment and seawater. Yatsu Tidal Flat is
considered to indicate succession from mud sedimenttideland to sandy tideland like other tidal flat in Tokyo Inner
Bay. Macrobenthic fauna is very diverse in sand tideland, so
the macrobenthic fauna in Yatsu Tidal Flat is considered to
become more diverse year by year (Fig.9). On the other hand,
Ulva spp. has become observed to be growing irregularly.
This large green algae is considered to beat macrobenthos
such as Annelida because of its large leaf body andsedimentation which leads anaerobic condition. From these
phenomena, ecosystem balance is considered to be not ingood condition [11].
Increase in the amount of pollutants that flows into Yatsu
Tidal Flat has been resulted from the concentration of
population, the increase of factories and the expansion ofreclaimed land. And water plant and macrobenthos,
especially Annelida in tideland is very useful to purify theorganic pollutants with their prey-predation interaction[12][13]. It is important to protect macrobenthos fromextermination because of its usefulness for purification oforganic polluted sediment, especially in tidal flat ecosystem.
In addition, Tokyo Inner Bay is recognized as a famous sea
area where the Blue Tide (Aoshio, anaerobic water bodies)occurs frequently. In fact, a large scale Aoshio occurred 3
times in summer in 2000. This anaerobic water bodies flew

Fig.8 Plankton observed in Yatsu Tidal Flat

Fig.9 Diversity index (H’) of macrobenthos

Fig.10 Sandbank located near the mouth of Yatsu river which
disappeared after the Tohoku Region Pacific Coast
Earthquake in 11 March 2011.
into Yatsu Tidal Flat and influenced greatly not only tomacrobenthos but also to other creatures in this enclosed tidal
flat. Furthermore, EDCs are also considered to be one ofenvironmental factors that give large effect [14]. So, further
study about the effect of environmental factors on
macrobenthic fauna is necessary for the environmental
improvement of not only Yatsu Tidal Flat but also any natureleft in urban area.

414

Fig.11 Relationship between geological and ecological succession in Yatsu Tidal Flat

6. EFFECT OF TOHOKU REGION PACIFIC COAST
EARTHQUAKEThe Tohoku region Pacific Coast Earthquake which
occurred at 11 March 2011 gave large influence not only tothe pacific coast but also to the coastal zone of Tokyo Inner
Bay. Yatsu Tidal Flat is not exception as same as Sanbanze
and Banzu Tidal Flat where also located in Chiba prefectureside of Tokyo Inner Bay. The liquefaction of sediment inYatsu Tidal Flat was observed. The promenadea walk wasdamaged and the microtopography of ebb water route in tidal
flat was changed with an uplift of sediment. For example, asandbank located near the mouth of Yatsu river wasdisappeared after the earthquake (Fig.10). It is considered thatmicro particle sand was ejected from underground andsedimental condition in the tidal flat became more sandy.
However, in Yatsu Tidal Flat, the water flow, especially ebb
tide, is very fast (about 2m/s in the canal) [15] and the sandparticle was immediately flush out to Tokyo Inner Bay. Ulva
spp. was also damaged temporary but it grew and occupied
the tidal flat surface again. Further continuous investigation
should be necessary to estimate the geological succession
effect on the tidal flat ecosystem structure and function(Fig.11).
7. CONCLUSIONS
The investigation of the effect of geological succession of
sediment on macrophyto and microbiota in aquifer ecosystem
was conducted in this study. The results obtained aresummarized as follows.
1) As a result of inflow loads has decreased withimprovement of public sewerage system, water quality haschanged from fresh water to sea water and sediment condition
has also changed from mud to sand, and bio-production
potential has reduced, in Yatsu Tidal Flat, the artificial
tideland lake.
2) There is the inherent clockwise flow occurred by tidal
action is made clear, and the relationship between theclockwise flow and sediment condition was suggested.
3) The phenomenon of sand spill is occurred by tide actionin the east area of Yatu Tidal Flat was made clear.

4) It is considered that Yatsu Tidal Flat has been changing
to the foreshore tidal flat again with remaining the
topographical character as tideland lake.
5) Because of the irregular growth of Ulva spp., thequantity of phytoplankton decreased drastically, and thedominant species of phytoplankton and zooplankton transitedfrom ideal to unsuitable food organisms for macrobenthos
which is important food for the migratory birds, and the
production potential of this aquifer ecosystem became infragility.
6) Further continuous investigation should be necessary toestimate the geological succession effect on the tidal flatecosystem structure and function.

8. REFERENCES[1]

Kurihara Y, “Ecology and Ecotechnology in Estuarine-Coastal Area,”
Tokai University Publishers, 1988.
[2] Yamashita H, “Ramsar Convention and Wetland in Japan,” Shinzansha
Scitech, 1993.
[3] Yatsuhigata Nature Observation Center, “Welcome to Yatsuhigata(tidal flat),” Narashino City, 1998.
[4] Murakami K, Ishii Y, Taki K, Hasegawa A, “Succession
Characteristics of Tideland Lake located in Tokyo Inner Bay,” J. of
Coastal Engineering, vol.47, Nov.2000, pp.1121-1125.
[5] Sugiyama K, “Sea Shore Biotope,” Shinzansha Scitech, 2000.
[6] Notoya M, “Utilization of Ulva spp. and Environmental Restoration,”
Seizando, 1999.
[7] Murakami K, Ishii T, Taki K, Ishii Y, Tatsumoto H, “Material
Circulation and Species Composition in Tidal Flat Microcosm
System,” Conference papers on 6th International Conference on the
Mediterranean Coastal Environment (MEDCOAST’03), vol.2,
Oct.2003, pp.825-830.
[8] Murakami K, Takano H, Katayama Y, Ogino Y, Mori T, Nagafuchi O,
Komai Y, Seiki T, “relationship between Macrobenthic Fauna andSediment Condition in the Seto Inland Sea,”
MEDCOAST99-EMECS99 Joint Conference, Land-Ocean
Interactions: Managing Coastal Ecosystems, vol.1, Oct.1999,
pp.111-122.
[9] Environmental Agency of Japan, Chiba Prefecture, Narashino City,“Environmental Research Report of Yatsu Tidal Flat,” 1996.
[10] Rhoads DC, Morse JW, “Evolutionary and Ecologic Significarse of
Oxygen deficient Marine Basins,” Lethaia, vol.4, 1971, pp.413-428.
[11] Ishii Y, Murakami K, Ishii T, Tatsumoto H, Taki K, “relationshipbetween Ecological Balance and Environmental Factors in Natural
Tidal Flat left from Reclamation,” Proc. Civil Eng. Ocean, JSCE,
vol.17, Oct.2001, pp.129-134[12] Inamori Y, Kimura M, Sudo R, “The Role of Benthos andConservational Measure in Tideland,” J. of Water and Waste, vol.36,
Jan.1994, pp.15-20.

415

[13] Kohama A, Enari K, Tamaki S, Nakayama M, “Evaluation of theAmount of Nitrogen and Phosphorus Absorption by the Aquatic Plant(Zizania latifolia),” J. of Japanese Society of Water Treatment Biology,
vol.39, Jun.2003, pp.59-66.
[14] Murakami K, Imatomi Y, Komai Y, Nagafuchi O, Seiki T, Koyama T,“Relationship between Annelida and Sedimental Environment in Seto
Inland Sea,” J. of Japanese Society of Water Environment, vol.21,
Nov.1998, pp.757-764.
[15] Yauchi E, Hayami T, Ishii Y, Tatsumoto H, “Environmental Dynamics
of Tokyo Bay Act on Yatsu Tidal Flat,” Environmental Information
Science. Extra, Papers on Environmental Information Science, vol.17,
Nov.2003, pp.327-330.

416

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

BioClean and Liquid Biofertilizers a New Way to the Green Area
Sittisak Uparivong, Department of Biotechnology, Faculty of Technology, Khon Kaen University, Thailand.

ABSTRACT: The one selection to solve the problems of severely
damaged by the tragic earthquake and tsunami is presenting a new
way to green area planet. The aim is to be creating a pilot project tosupport green area for environmental places living. BioClean andliquid biofertilizer projects are attempting to be a new way for thenatural balance of combined green area. This generation of bioclean
had produced by the donation of various flowers during thegraduated ceremony of khon kaen university since 2010, and adding
the 18 zymogenic synthetic microorganisms (18 ZSMs) withmolasses as substrate. Furthermore, the liquid biofertilizers had
produced as similar processes as bioclean excepting differential raw
materials by using the previous research product of liquidbiofertilizers, which were produced by using vegetables, fruits,
herbal crops, etc. Results of microorganism population of bioclean
and liquid biofertilizer products were 1.0 x 108 to 8.0 x 1012 cfu/ml,
which mainly serviced for environmental and agricultural sections,
respectively. Including, the both products were presented focus on
supporting to the green area for living places and related to climatechange. The service result of project products was activated on theevent of “Temples Big Cleaning Day 2011” by servicing forcleaning and the green area at the two temples near khon kaen
university, which satisfactory result obtained 50% excellence, 30%
good and 20% fair by the board of ten committee’s considerations.

Keywords: BioClean, liquid biofertilizer, 18 zymogenic
synthetic microorganism (18 ZSMs), temple, green area
1. INTRODUCTION
BioClean is an original brand name of liquid bioextracted
flowers since 2003 [10]. BioClean was produced as thesimilar processes as the production of liquid biofertilizers byfermenting various biomaterials. The difference between
bioclean and liquid biofertilizers was the differential raw
materials, which liquid biofertilizers were produced by using
various combined biomaterials such as; vegetables, fruits,
some herbaceous crops, etc., while bioclean was produced byusing the diversities of odoriferous flowers (donated various
flowers from a graduated ceremony of khon kaen university
since 2010) in order to solving the odour problem of liquidbiofertilizers that responded by users. The target project of
this generation of bioclean was mainly used to service for theexample communities of environmental and healthy living to
the two temples of adulkaewmordee temple and nonmuang
temple at nearly to the regions of khon kaen university inorder to servicing for reducing odour treatment such as; onfloor temples, sewerage, bathroom/lavatory, wastewater
treatment, water quality treatment etc., and liquid biofertilizer
services were mainly used to encourage for green area andalong the main road of khon kaen university. Furthermore,
bioclean could be applied for agricultural sectors as thesimilar as liquid biofertilizers. Thus, the utilization of liquidbiofertilizers were to solve the problem of decay soil byincreasing organic biofertilizers into soil land, and to increase
quality organic agricultural productions [1], and to quality
better life and health safety, and to respond the policy for
agricultural products safety to kitchen of the world [3]. The
board of 18 zymogenic synthetic microorganisms (18 ZSMs)
were consisted of 9 bacteria, 6 fungi and 3 yeast species. The
18 ZSMs was originally selected by Prof.Dr.Chaitat Pairintra
at khon kaen university to our research project of“Improvement of the Theparuge’s Liquid Biofertilizer
Product” since 2002 [2].
2. MATERIALS AND METHODS
2.1 Materials
The raw materials were used to produce this generation
products as 1). To produce bioclean (version 2011) was
utilized the donated various flowers during the graduatedceremony of khon kaen university in 2010. 2). To produce
liquid biofertilizers (version 2011) were utilized the previous
research stocks of original microorganism seeding of liquidbiofertilizer biotech-1 and liquid biofertilizer biotech-2 which
had been produced by using organic biomaterials such as;vegetables, fruits, herbaceous crops, etc. since 2004-2005[3]-[5]. 3). The original microorganism seeding of liquidbiofertilizer biotech-1 or liquid biofertilizer biotech-2 were
produced by the group of 18 zymogenic synthetic
microorganisms (18 ZSMs) that was transferred from themicroorganism seeding product of a liquid biofertilizer
KKU-1 [2]. 4). Molasses was supplied from a sugar canefactory at the local area of udon thani near khon kaenprovinces.
2.2 Equipments
The biofermentor equipments utilized to produce thisgenerated products of bioclean and liquid biofertilizers were
supported by the previous research equipments such as; a300-L biofermentor model BT-1 equipped with a stirrer
motor ½ hp [2], a 500-L liquid biofermentor model BT-1
equipped with a motor 1 hp and/with/or/without a compressor
air supply [3], a 500-L liquid biofermentor model BT-2
equipped with a stirrer motor 1 hp and/with/or/without an air
pump supply [5], and a 500-L liquid biofermentor model
BT-3 equipped with a stirrer motor 1 hp and/with/or/without
an air pump supply [7], as shown in Fig.1.
2.3 Methods
The method to produce bioclean was produced as the sameprocesses method as the production of liquid biofertilizers but
excepting only the differential raw materials [8]-[11]. This
generation of bioclean (2011) was used variously donated417

Fig.1 Liquid biofermentors.
flowers as mention above while liquid biofertilizers were
utilized the previous research stocks of original liquidbio-extract of vegetables, fruits, some herbaceous crops, etc.,
the 3 step processes methods as;
The first step of preparation of original liquid bio-extract(OLB), the all raw materials of variously donated flowers
were cleaned and cut into small pieces for fermenting as theratio as “small pieced flowers : molasses : liquid biofertilizer
microorganism seeding (LBMS) of liquid biofertilizer
biotech-1 or biotech-2” = 3 : 1 : 1 or 3 : 1 : 2 (w/v), and cleanwater within biofermentors such as; 75.7, 113.55, 151.4 liters(20, 30, 40 gallons) or more etc., during the retention time of
1 week.
The second step of fermentation, the LBMS product fromthe first step process was fermented with molasses and cleanwater as the ratio as “LBMS : molasses : clean water” = 1 : 1 :40 or 2 : 1 : 40 (v/v) within liquid biofermentors such as; 500,
1000, 1500 liters equipped with/or/without a stirrer motor to
produce the fermented liquid bioproduct during the retention
time of 2 weeks.
The third step of filtration, the fermented liquid bioproduct(FLB) from the second step process was filtered to obtain the
final products as so called “BioClean” or “Liquid
Biofertilizers” depending on the differential types of raw
materials.
The methods to utilization of bioclean and liquidbiofertilizers (including 18 ZSMs) are aimed to the two wayof utilization for environmental and agricultural sections as;1). For agriculture, using the dilution ratio of liquidbiofertilizer product at 1:2000 by spraying or pouring to thegrowth crops every 5-7 days, and the dilution ratio of 1:500 to
the plant trees and green area for lively places.
2). For environment, using the dilution ratio of bioclean
product at 0.05% (1:2000) for floor cleaning, and theconcentration ratio of 70-80% for reducing odour treatmentof bathroom/ lavatory/ toilet/ wc./ sewerage etc., and theconcentration ratio of 0.05% for wastewater treatment andwater quality treatment.
The methods to quality testing of bioclean and liquidbiofertilizers were investigated before servicing to thetemples and the green area at the regions of khon kaen
university as; 1). For agricultural testing, the both products
were tested by cultivating for various crops such as; waterconvolvulus, flowers, etc., 2). For environmental and healthy
testing, the both products were tested by treating for sanitarysystems such as; reducing odour treatment for bathroom/
lavatory/toilet/wc., sewerage, wastewater treatment and water
treatment, including water quality treatment for goldfish andnile tilapiafish living, etc.
2.4 Analysis Methods
The analysis composition of bioclean and liquid biofertilizer
products were supported investigation by a laboratory of
faculty of agriculture of khon kaen university, and referred tothe previous analysis methods of the composition of liquidbiofertilizers such as; pH, EC, %OM, N, P, K, Na, Ca, Mg,
etc., as in [2]-[7]. Microorganism biomass populations were
determined by the method of standard plate count (agarpowder, peptone, bacteriological HIMEDIA RM001). The
quality standard of both products were determined by impact
testing for agricultural, environmental and healthy impact
assessments. For agricultural testing, the quality of bothproducts were evaluated the growth rate of crops by pots/
fields testing such as; water convolvulus, some flower, etc.
For environmental and healthy impact assessments, thequality of both products were evaluated by field testing at
sanitary systems such as; reducing odour treatment forbathroom/lavatory/toilet/wc./sewerage at general households
etc., and quality testing for the efficiency of water treatment
such as; DO, BOD, COD, TKN, etc., at a pond treatment of
nonmuang temple near khon kean university.
The analysis methods to evaluate the satisfactory services
rating of both products of bioclean and liquid biofertilizers
during temples big cleaning day at wat adulkaewmordee andwat nonmuang, and green areas and along main road at theregion of khon kaen university, were evaluated by surveying
the applied checklist forms of academic service center ofkhon kaen university, the evaluated methods as following; 1).
Evaluation of satisfactory services rating by general persons/home/shop owner/or/members were evaluated by surveying
at around the regions of the both temples. 2). Evaluation ofsatisfactory services rating by persons and monks were
evaluated by surveying at the both temples. 3). Evaluation ofsatisfactory services rating by the board of expert committees
were evaluated by inviting the expert board of ten
committee’s considerations.

418

3 RESULTS AND DISCUSSION
3.1 Properties products of BioClean and Liquid
Biofertilizers
The products of bioclean and liquid biofertilizers of biotech-1
and biotech-2 (2011) were contained into bottom and tankcontainers such as; 1-L, 10-L, 20-L, etc., as shown in Fig. 2.
The results of effective microorganism populations of
bioclean and liquid biofertilizers biotech-1, 2 obtained 1.0 x
108 to 4.0 x 1012 cfu/ml and 1.0 x 108 to 8.0 x 1012 cfu/ml after
the retention time more than 7-8 day respectively, which be
more than the standard products (107-108 cfu/ml), theproperties of both products obtained such as; pH = 3.76, EC =4.31 ds/m, N = 0.018 ppm, P = 25 ppm, K = 561 ppm, Na = 73
ppm, Ca = 254 ppm and Mg = 200 ppm for bioclean, and pH= 3.45-4.19, EC = 3.00-5.43 ds/m, N = 0.025-14 ppm, P =28-38 ppm, K = 881-1023 ppm, Na = 108-225 ppm, Ca =175-271 ppm and Mg = 0.50-142 ppm for liquid biofertilizers
biotech-1, 2, as shown in Table 1, and as similar results as theprevious research products of bioclean and liquidbiofertilizers, as in [2]-[11].

Fig. 2 BioClean and Liquid biofertilizer biotech-2 (2011).

Table 1
Properties products of bioclean and liquid biofertilizers biotech-1, 2.
Product Sample(2011)apHw

EC(ds/m)

2.50
BioClean (donated flowers)
4.32 (1:5)b
4.31
BioClean
3.76
3.00
Liquid biofertilizer Biotech-1 4.19 (1:5)b
3.45
5.43
Liquid biofertilizer Biotech-2
a
The generational products were produced since 2011.
b
The pHw (1:5).
3.2 Quality Testings of Products
The quality products of bioclean and liquid biofertilizers
biotech-1, 2 were satisfyingly evaluated by quality testing
before distributing to general users at the pots/fields tests. For
agricultural testing, the both products of bioclean and liquidbiofertilizers were satisfactory the growth rate of various
crops such as; water convolvulus, flower, etc. during 45 days[2]-[11]. For environmental and healthy testing, the bothproducts were satisfactorily evaluated by testing for sanitarysystems such as; reducing odour treatment for the bathroom/
lavatory/ toilet/ wc./ sewerage of distributed generalhouseholds by responding feedback users, and the efficiency
bioclean of pond treatment was 61.90% and 89.80% during
the retention time of 7 day and 14 day, respectively at a pond
treatment of wat non muang near khon kaen university,
including water quality treatment by using the dilution ratioof 0.05% bioclean for goldfish and nilefish healthy living.

OM(%)

Total N(ppm)

Total P(ppm)

Total K
(ppm)

1.43

10
0.018
14
0.025

26
25
28
38

598
561
811
1023

Total
Na
(ppm)
82
73
108
225

Total
Ca
(ppm)
205
254
271
175

Total
Mg
(ppm)
131
200
142
0.50

results which obtained 44.4% excellence, 33.3% good, 22.2%
fair for wat adulkaewmordee, and 50% excellence, 30% good,
20% fair for wat nonmuang after services respectively, which
considered by invited the expert board of ten committee’s
considerations.

3.3 Services Rating of Temples Big Cleaning Day
The both products of bioclean and liquid biofertilizers wereserviced to the two temples of wat adul kaew mordee and wat
non muang near the regions of khon kaen university, as
shown in Fig. 3, on the event of “Temples big cleaning day”
by using the dilution ratio of bioclean at 1:2000 for floor
cleaning, 70-80% for thin coating sewerage/toilet/wc., 0.05%
for wastewater treatment or water quality treatment, andusing the dilution ratio of liquid biofertilizer biotech-1, 2 at
1:500 for the temples plant trees and green area. The
overview services to the both temples were satisfactory
419

Fig. 3 The Two Temples Big Cleaning Services.

3.4 Services Products to Green Area
The utilization of liquid biofertilizers biotech-1, 2 wasserviced in order to being liquid biofertilizers for the green
area living and the plant trees along the main road at theregions of khon kaen university by using at the dilution ratioof 1:500, as shown in Fig. 4. The overview services of liquidbiofertilizer products were satisfactory for green area living
place during May-August, 2011.

Fig. 4 Service of liquid biofertilizer biotech -1 (v.2011) atalong the main kku road to the green area living.

4 CONCLUSION
1). BioClean services to the temples
BioClean (v.2011) services from the board of 18 zymogenic
synthetic microorganisms (18 ZSMs) could be reduced odour
treatment for sanitary systems and communities healthyliving places to the pilot temples with satisfactory services.
2). Liquid biofertilizers services to the along main road andthe green area
Liquid biofertilizers biotech-1, 2 (v.2011) services from theboard of 18 zymogenic synthetic microorganisms (18 ZSMs)
could be much more enriched to the plentiful trees of along
main road and green area for environmental places living andrelated to climate change with satisfactory services.

[3]

Uparivong, S. et al. 2004. Production of liquid biofertilizer to kitchen of
the world. In Proc. The 16th Annual Thai Society for Biotechnology“Innovative Biotechnology: The Opportunity for Kitchen of the World”.
Phitsanulok, 12-15 December.
[4] Uparivong, S. 2005. The production of liquid biofertilizer. In Proc.
Asian Wetland Symposium 2005. Innovative Approaches to
Sustainable Livelihood. Bhubaneswar, India, 6-9th February.
[5] Uparivong, S. et al. 2006. Liquid biofertilizer fermentor andmicroorganism biofermentor of model BT-2. In Proc. Int. Conf. on
TISD2006 Technology and Innovation for Sustainable Development
Conference. Khon Kaen, Thailand, 25-26 January.
[6] Uparivong, S. 2006. 18 Zymogenic synthetic microorganism of liquidbiofertilizer Biotech-3 for agricultural and environmental. In Proc. The
18th Annual Thai Society for Biotechnology. Biotechnology : Benefits& Bioethics. Bangkok, Thailand. 2-3 November.
[7] Uparivong, S. 2007. The development of biofermentor BT-3. In Proc.
IAEC2007 International Agricultural Engineering Conference. Theme :Cutting edge technologies and innovations on sustainable resources forworld food sufficiency. Asian Institute of Technology (AIT), Bangkok,
Thailand, 3-6 December.
[8] Uparivong, S. 2009. BioClean. In Proc. Int. Conf. on Innovations inAgricultural, Food and Renewable Energy Productions for Mankind.
The 10th Annual Conference of Thai Society of AgriculturalEngineering. 1st-3rd April 2009. Surasammanakhan, Suranaree
University of Technology, Nakhon Ratchasima. pp. 173-175.
[9] Uparivong, S. 2009. Utilization of liquid bioferilizer for mandkinds. In
Proc. XXXIII CIOSTA CIGRV Conference 2009. Technology andManagement to Ensure Sustainable Agriculture, Agro Systems,
Forestry and Safety. WORKSHOP IUFRO. Forestry Utilization in
Mediterranean Countries with Particularly Respect to Sloping Areas.
17-19 June 2009. Reggio Calabria, Italy. pp. 1391-1395.
[10] Uparivong, S. 2010. BioClean a new friendly environmental cleaningproduct. In Proc. Int. Conf. Workshop on “Livelihood and HealthImpacts of the Climate Change: Community Adaptation Strategies”
24-25 August 2010 Pullman Raja Orchid Hotel, Khon Kaen, Thailand.
pp. 122-128.
[11] Uparivong, S. 2010. BioClean for environmental biotechnology ofcanteens and green area. In Proc. Int. Conf. on Biotechnology for
Healthy Living. The 22nd Annual Meeting of the Thai Society for
Biotechnology. TSB2010 October 20-22, 2010. Prince of Songkla
University, Trang Campus, Trang Province, Thailand. pp. 849-857.

5 ACKNOWLEDGMENT
This research was financially supported by grants from theacademic service center of khon kaen university, Thailand.
Acknowledgements are extended to the expert board of ten
committee’s considerations to Mr. Boonyarit Sombutlai, Mr.
Rachaporn Patawanit, Mr. Soontorn Sarapanich, Asst. Prof.
Dr. Araya Ahromrit, Asst. Prof. Dr. Somsak Pitaksanurat,
Assoc. Prof. Dr. Sirinda Yunchalard, and Assoc. Prof. Dr.
Pinit Wangsomnuk, Pol. Sub. Lt. Suwith Nampan, Pra
Petcharat Kachapanyo, Pra Veerasak Jantemduang.
6 REFERENCES[1][2]

Pairintra, C. et al. 1991. Zymogenic synthetic soil and crop
“Agricultural in the Next New Century”. ISBN 974-55-251-3.
Uparivong, S., Prachankanchana, S. and Pairintra, C. 2002. Productdevelopment on liquid biofertilizer of the theparuge’s agriculturalcommunity. J. of Academic Service Center Khon Kaen University. 11(3): 37-41.

420

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Massive Point Cloud Acquisition on a Soil Surface Using a Stereo Camera
Masafumi NAKAGAWA, Anna NAKANISHI, Hirotaka ENDO and Kenta OCHIAI,

Department of Civil Engineering, College of Engineering, Shibaura Institute of Technology


ABSTRACT: Point cloud data can describe natural objects such assoil. Massive point cloud data acquisition is required for geographic
information systems (GISs) and remote sensing to generateenvironment maps, including surface models. For the contactlessmeasurement of landslides, we believe that a stereo system has someadvantages, because it can acquire high spatial resolution data in realtime. Therefore, we propose massive point cloud acquisition fromsoil surfaces using a stereo camera without markers. Although soil
surfaces are difficult candidates for image processing, we haveverified in indoor experiments simulating landslides that our
procedure with constraints can measure a soil surface using a stereo
camera. We have also tried to develop a temporal stereo
measurement application based on the above procedure. Although
our experiments were conducted indoors, we have confirmed thatour procedure can be used as a 3D measurement system withoutmarkers for moving soil surfaces.

Keywords: Landslide, Stereo measurements, Image matching,
Point cloud, Point-based rendering
1. INTRODUCTION
A set of points with geometric X, Y, and Z values is called
point cloud data. When point cloud data are registered with acolor image taken from a digital camera, each point in the
point cloud can hold color information. Because the data canbe represented without topological relationships between
points, the point cloud can describe natural objects such as
soil, rubble and trees. Massive point cloud data acquisition isrequired for geographic information systems (GISs) andremote sensing to generate environment maps, including
surface models and elevation models. Developers ofautonomous construction robots are also studying point clouddata acquisition in outdoor environments [1].
Stereo cameras and laser scanners are general survey
instruments for acquiring point cloud data in an outdoor
environment. The stereo camera can acquire massive point
clouds using consumer cameras. Data from a scene can beacquired at 15 or 30 frames per second. However,measurements depend on the density of available featurepoints in an image. Soil surfaces are difficult to measurecorrectly, because there are generally few feature points forimage processing on the surface. On the other hand, a laser
scanner can acquire massive point clouds and does not requirefeature points. A current high-end scanner can acquireapproximately one hundred thousand points per second.
However, data acquisition for one scene requires from a few
seconds to a few minutes. Thus, in the current state, there are
no available 3D measurement systems without markers formoving soil surfaces.

We believe that a stereo system has advantages for soil
surface measurement, because it can acquire data in real time.
Therefore, we propose massive point cloud acquisition fromsoil surfaces with homogeneous textures using a stereo
camera without markers. In addition, we try to develop a
temporal stereo measurement application to represent
landslides.
2. METHODOLOGY
In general, when massive point cloud data are acquired with astereo camera, long processing times are required. In addition,
when the measured object is a soil surface, the homogeneous
textures in images cause high error rates because of
mismatching in the stereo matching step. When landslide
animation is generated with temporal point cloud data, 3D
soil motion representation is unstable because of the above
errors. On the other hand, automated point cloud modification
can be achieved with hole-filling approaches [2, 3, 4],although they require long times for massive point processing
and error point detection before the modification. Generally,
error point detection also requires long processing times.
Therefore, we focus on performance improvement to reduce
processing time and errors. First, we propose a stable stereo
matching procedure based on a coarse-to-fine approach to
reduce processing time and mismatching in stereo matching.
Second, we propose a temporal stereo matching methodology
using some constraints to improve stability in tracking
moving surfaces.
2.1 Massive point cloud acquisition from a soil surface
Figure 1 shows our procedure to acquire massive point cloudfrom soil surfaces. It consists of three components: edge
density image generation for feature point distribution;correlation coefficient matching with an epipolar constraintusing a circular template; and coarse-to-fine image matching
using a spatial constraint with iterative surface model
generation.
Stereo images
Rectified images

Calibration
parameters
Edge density image

Mask image
Feature points
Surface fitting
Initial DSM
Coarse-to-fineDSM generation

Correlation matching
using circle template

Stereo matching using
initial DSM
Massive point cloudFig. 1 Stereo matching with coarse-to-fine DSM generation421

We extract points with high likelihoods of features from abase stereo image to achieve an accurate matching, because
the use of suitable points for image matching can improvematching success rate. First, an edge density image is
generated from an edge image taken from the base image.
Each pixel in this edge density image expresses thecomplexity of an edge. Then, feature points with higher edge
densities are selected from the edge density image as suitablefeature points to generate the initial digital surface model(DSM).
In coarse-to-fine matching in image space, fine matching
generally uses higher-resolution stereo images and initialvalues generated with lower-resolution stereo images [5].
This procedure achieves faster processing and stablematching. However, when there are only a few suitable points
for features in an image (e.g., a sand surface), unsuitable
points selected for image matching cause many matchingerrors. Therefore, in this research, coarse-to-fine matching in
3D space is applied to stereo matching, as shown in Fig. 2.
First, 3D coordinate values are acquired with correlation
matching using only feature points extracted from the edge
density image. In our procedure, the sum of absolutedifference (SAD) methodology is applied to correlation
matching. Second, coarse 3D shaped data are generated as an
initial DSM with spatial interpolation such as bilinear
interpolation or cubic spline interpolation. Then,corresponding full points in the stereo reference image aresearched for under the DSM constraint, as described later.
Feature points P (XI, YI, ZI)
Grid points PI (X,Y,Z)

Search range withtraditional approaches(= Wide area)(Interpolated points)

If ARt1 is different from BRt1, this result can be rejected as amatching error.
In addition, three constraints are implemented to reducemismatching. The first constraint is an epipolar constraint.
When stereo images are rectified using camera calibration
parameters, the search range can be limited to the horizontal
direction of feature points. The second constraint is the DSM
constraint, which makes the search range narrower along the
epipolar line. The third constraint is a motion constraint. This
constraint depends on objects. For example, when a sand flowsuch as a landslide is an object, a search region in the image
can be limited to downward from the initial feature point,
because the sand flow moves downward.
Stereo image
Time

Left image

Right image
DSM constraint

Feature pointt0

Search range
Stereo matching
point ARt0
EpipolarPrevious position

Previous position constraint

Motion constraint

Motion constraint

Search range

Search range

t1

Search range
DSM constraint
Image matching
point B Lt1

Matching result Matching result
A Rt1
B Rt1

Fig. 3 Our matching methodology with mismatching
rejection using temporal stereo images; ARt0, ARt1, BLt1 andBRt1 indicate image coordinate values (i.e., ALt1 is an image
coordinate value in the left image at time t1)

Search range in our approach(= Narrower search range)

3 EXPERIMENTf（x,y）

Initial DSM

We conducted an indoor experiment, as shown in Fig. 4. We
first prepared a stereo camera consisting of two digital
cameras controlled by a PC. These cameras were assembled
and calibrated to satisfy the required measurement accuracy,
as shown in Table. 1.

Initial DSM

f(x,y) surface

Full-point matching(x,y,z) points

Right images

Fig. 2 Coarse-to-fine DSM generation

0/30[sec]

2.2 Moving soil surface tracking
The tracking of a moving soil surface is processed using
temporal stereo image matching in the two steps of stereo
matching and temporal image matching. Our temporal stereo
matching consists of a clockwise matching flow and an
anticlockwise matching flow, as shown in Fig. 3. Temporal
stereo images extracted at times t0 and t1 are arranged in a
vertical direction in the figure. Image coordinate values x andy are translated to A* or B*.
In the clockwise matching flow, the corresponded point (ARt1)
is detected in the right image at time t1 after image matchingin the right image at time t0. In the anticlockwise matching
flow, the corresponded point (BRt1) is detected in the rightimage at time t1 after image matching in the left image at time
t1. When ARt1 is equal to BRt1, image matching is successful.
422

1/30[sec]

2/30[sec]

3/30[sec]

4/30[sec]

Fig. 4 Experimental environment

Fig. 5 Acquiredsequence images

Table. 1 Stereo camera specification
CameraNikon D300

Focus length
Baseline

35 mm
15 cm

Distance from camera to object
Measurement accuracy
Image size (still capture mode)
Image size (30fps mode)

150 cm0.20 mm(RMSE)
1072×712 pixel
640×426 pixel

Finally, Fig. 8 shows a result of temporal image matching
using 58 frames taken at 30 frames per second on the left
image coordinates. The blue line shows a tracking result
using 390 distributed feature points on the moving sandsurface.

We also prepared sand of approximately 1 mm particle size ina tray as the measured object. The stereo camera thencaptured this sand object. In this operation, we inclined thesand tray from 30° to 45° to create a sand flow to simulate alandslide. Sequenced stereo images were acquired at 30
frames per second using the stereo camera, as shown in Fig. 5.

y[pix]

x[pix]

Fig. 8 Result of temporal image matching

4 RESULTS
Feature extraction, 3D measurement and temporal image
matching were conducted in our experiment for massive point
cloud acquisition on a soil surface using a stereo camera. First,
a result for feature extraction in the initial DSM generation isshown in Fig. 6. The left image shows a color image used as
input data, the center image shows an edge density image
taken from a region selected manually in the input data, andthe right image shows the result of the masking procedure on
the edge density image. These results show that feature points
can be extracted from higher edge density values indicated as
white areas in the edge density image.

Fig. 6 Result for feature extraction for initial DSM generation
Second, 3D measurement results after stereo matching areshown in Fig. 7. The left image shows a result of the
traditional approach without coarse-to-fine DSM generationin stereo matching, and the right image shows a result with
our methodology.

5 DISCUSSION
The stereo measurement accuracy was 0.20 mm, as describedabove. Thus, our approach should be considered in terms of
performance improvement in matching success rate andprocessing time. In addition, corresponding feature points
from images selected manually are hard to use as true values
for validation, because it is difficult to find objects manuallyin soil surfaces with homogeneous textures. Therefore, some
performance comparisons of our approach with traditional
approaches are included in our validations.
Although soil surfaces with homogeneous textures are alsodifficult candidates for image processing, we have verified
that our procedure with constraints can measure a soil surface
using a stereo camera in indoor experiments simulatinglandslides, as follows.
First, we have verified the accuracy of the matchingprocedure. Figure 9 shows a comparison of our approach withthe traditional approach in 3D measurement results projected
onto the Z-plane. The left image shows the result for the
traditional approach, and the right image shows the result for
our approach. With the traditional approach, the result
includes some matching errors, shown as holes in the left
image. On the other hand, the results for our approach show
no holes, as shown in the right image.

Fig. 9 Comparison of our approach with the traditional
approach for 3D measurement results projected onto theZ-planeFig. 7 Comparison of our approach with the traditional
approach for 3D measurement results

We have subtracted the results from our approach from theresults of the traditional approach, and calculated absolute
subtracted values to measure the matching error, as shown in423

Fig. 10. The subtracted values in 3D space are colored asmatching error values with an index from 0 mm to 60 mm; thewhite region indicates a matching error of more than 60 mm.
The results can clearly be classified as correctly matched or
mismatched.
In addition, we have counted the pixels taken from the region
selected in the input image to compare our approach with thetraditional approach as errors. The number of pixels takenfrom the region selected in the input image was 279,818
points. Of these, 181,720 points are within a 5 mm gap, and196,347 points are within a 10 mm gap.
This numerical comparison indicates that the matching
success rate for the traditional approach in 3D data generation
is from 60% to 70%, while our approach can reducemismatching error by between 30% and 40%.

60
50

Z [mm]

X [mm]

Y [mm]

X [mm]

Y [mm]

40

6 CONCLUSION30

We have proposed massive point cloud acquisition from soil
surfaces with homogeneous textures using a stereo camera
without markers. Although soil surfaces are difficult
candidates for image processing, we have verified that our
three-component procedure can measure a soil surface using
a stereo camera in indoor experiments simulating landslides.
We have also developed a temporal stereo measurement
application based on the above procedure. Although ourexperiments were conducted indoors, we have confirmed thatour procedure can be used as a 3D measurement system
without markers for moving soil surfaces.

20
10Fig. 10 Subtracted 3D measurement result projected onto theZ-planeSecond, we have compared the processing time for ourapproach with that for the traditional approach, as shown in
Table. 2. The search range could be narrowed from 800 mm
to 100 mm in our experiment. As a result, although theperformance comparison depends on the parameter settings,
such as the range for matching, Table. 2 shows that ourapproach performed about 10 times faster than the traditional
approach.
Table. 2 Processing time comparison of our approach with thetraditional approach in processing time
Traditional approach

Our approach

Stereo matching fromZ = 1100 mm to Z = 1900 mm

Stereo matching withinitial DSM ± 50 mm

Edge density image generation

-----

25 s

Initial DSM generation

-----

3s

Stereo matching (Sub-pixel matching)2247 s195 s2247 s

223 s

Total

Z [mm]

Fig. 11 Comparison of our approach with the traditional
approach for 3D feature tracking results of moving sandsurface shown as blue lines; the left image is the result for the
traditional approach, while the right image is the result for ourapproach

Subtracted values
in 3D space[mm]Measured points：280,000
Intel Core i7 2.80 GHz／Single-core／MATLAB

the moving sand surface in 58 frames, and the Z-axis is thedepth direction from the stereo camera. We have verified thatmatching errors given by the traditional approach have been
rejected automatically with our approach.

Third, we have compared the performance for our approachwith that for the traditional approach for temporal stereo
measurement and tracking. The 3D feature tracking results
for the moving sand surface are shown in Fig. 11. The bluelines indicate the 3D feature tracking results of 390 points on

7 REFERENCES[1] Shiegru Sarata, Niriho Kiyachi, Kazuhiro Sugawara,“Measuring and Update of Shape of Pile for Loading Operation
by Wheel Loader”, The 25th International Symposium on
Automation and Robotics in Construction. ISARC, pp.
113–118, 2008.
[2] Wei, Z. Shuming, Gao. Hongwei, Lin., “A robust hole-fillingalgorithm for triangular mesh”, The Visual Computer, 23(12):987-997, 2007.
[3] Zoltan, C.M. Radu, B.R. Michael, B., “On Fast Surface
Reconstruction Methods for Large and Noisy Point”, Robotics
and Automation, ICRA '09, IEEE International Conference:
3218 -3223, 2009.
[4] Lars, L., Karsten, M. & Paul R., “Splat-based Ray Tracing ofPoint Clouds”, Journal of WSCG, Vol.15, Issue: 1-3: 51-58,
2007.
[5] Masafumi NAKAGAWA, Ryosuke SHIBASAKI,“Development of Methodology for Refining Coarse 3D Urban
Data Using TLS Imagery”, International Society for
Photogrammetry and Remote Sensing, 2004.

424

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Estimation of Earthquake Ground Motion in Padang, Indonesia
R.R. Rusnardi. J. Kiyono. Y. Ono., Graduate School of Engineering, Kyoto University, Japan., Department
of Urban Social Systems and Civil Engineering, Tottori University, JapanABSTRACT: Several powerful earthquakes have struck Padang
during recent years, one of the largest of which was an M 7.6 eventthat occurred on September 30, 2009 and caused more than 1000
casualties. Following the event, we conducted a questionnaire
survey to estimate the shaking intensity distribution during theearthquake. About 500 residents of Padang were interviewed. The
residents received explanations for each item on the questionnairefrom the interviewers, and answers were filled in directly on theanswer sheets. From this survey we produced a map of the shaking
intensity distribution in Padang. In addition to the questionnaire
survey, we performed single observations of microtremors at 110
sites in Padang. The results enabled us to estimate the site-dependent
amplification characteristics of earthquake ground-motion. We also
conducted a 12-site microtremor array investigation to gain arepresentative determination of the soil condition of subsurface
structures in Padang. From the dispersion curve of arrayobservations, the central business district of Padang corresponds torelatively soft soil condition with Vs30 less than 400 m/s, thepredominant periods due to horizontal vertical ratios (HVSRs) are in
the range of 2.0 to 4.0 s, and the seismic intensity obtained is upper 5(5+) in the JMAi scale. By making these observations, we can obtaina relationship between soil types, predominant periods and seismic
intensities.
Keywords: peak ground acceleration, seismic intensity of the Japan
Meteorology Agency, Padang Earthquake, microtremor
observations.

2009 (Mw 7.6). The hypocenter of the Padang earthquakethat occurred on September 30, 2009 was located in the oceanslab of the Indo-Australian Plate at -0.81°S, 99.65°E and at a
depth of 80 km. It produced a high degree of shaking and thetremor was felt in the Indonesian capital, Jakarta, about
923 km from the epicenter. The tremors also were felt inneighboring countries such as Malaysia and Singapore [2].
The earthquake caused landslides and collateral debris flows
in the hills surrounding Lake Maninjau. A major landslide in
Gunung Nan Tigo, Padang Pariaman completely destroyedsome villages and forced road closures.
This 1900-km-long active strike-slip fault zone that runs
along the backbone of Sumatra poses seismic and fault
hazards to a dense population distributed on and around thefault zones [3]. The Sumatran Fault is highly segmented. It
consists of 20 major geometrically defined segments and theslip rate along the fault increase to the northwest, from about
5 mm/yr [3]. This fault also has generated large destructive
earthquakes, e.g., 1892 (Mw 7.1), 1943 (Mw 7.6) and 2007
(Mw 6.4). These faults are capable of generating strong
ground motion in the future that would greatly affectvulnerable structures. According to our catalogs, theSumatran Fault produces a very high annual rate ofearthquakes, many of which occur in the shallow region
under the island of Sumatra (Fig. 1).

1. INTRODUCTION
Seismic intensity is one of the simplest and most importantparameters for describing the degree of ground shaking
during an earthquake. Sometimes it has a strong correlation
with the human response to ground shaking, observations ofdamage, and earthquake effect. In this study, we adopt theseismic intensity scale of the Japan Meteorology Agency(IJMA) scale, an instrumental seismic intensity that was first
adopted in Japan after the 1995 Kobe earthquake. Its scaleruns from 0 to 7, with ten classes including lower 5, upper 5,lower 6 and upper 6. We propose the use of the instrumental
IJMA to make initial and fast estimations of damage following
a seismic event.
The city of Padang, located on the west coast of Sumatra inwestern Indonesia, lies close to the Sumatran subduction zonethat is formed by the subduction of the Indo-Australian Platebeneath the Eurasian Plate. Relative motion of the plates
occurs at a rate of about 50 to 70 mm/year and this is the main
source of subduction-related seismicity in the area [1]. Basedon our catalog, seven giant earthquakes have occurred in thisregion since records began: 1779 (Mw 8.4), 1833 (Mw 9.2),
1861 (Mw 8.3), 2004 (Mw 9.2), 2007 (Mw 7.9 and 8.4) andLegendMw>6.5<100km depth

Fig.1 Seismicity of Sumatra Island from 2005 to 2010,
Mw>6.5, <100km depth, and Padang city425

2. REGIONAL GEOLOGY AND RECENT
EARTHQUAKES
The city of Padang, with a population of 856,814 people as of2008, is the capital of West Sumatra province. The location of
the city center is at 100.38°E, 0.95°S. The main part ofPadang is situated on an alluvial plain between the Indian
Ocean and the mountains. For the most part, the mountainous
area is formed of Tertiary sedimentary rocks with outcrops ofmetamorphic rocks seen in some places. The alluvial plain
spreads along the base of the mountains and is roughly 10 km
wide in the east-west direction and 20 km wide in thenorth-south direction.
The topography of the Padang region (Fig.4(a)) is very
similar to the tsunami-damaged area of Miyagi Prefecture in
Japan, that was inundated by as much as 4-5 km from thecoast after the March 11, 2011 Mw 9.0 Tohoku earthquake
off the east coast of Honshu. In Padang, about 600,000 people
live in the coastal area (covering about 60 km2). The
population density is very high, about 8500 people/km2. The
city is located on the coast of the Indian Ocean between theSumatran Fault and the Sunda Trench Fault. Both faults areactive with slip rate ranging from 10 to 27 mm/year [3].
According to our catalog, 2995 events with a magnitude
greater than 4 occurred in this region from AD 1779 to 2010.
The seven giant earthquakes mentioned previously have all
been strongly felt here. For example, the source of the 2009
Padang earthquake was located in the ocean slab of the
Indo-Australian Plate.
It produced extensive shaking and severe damage tohouses and buildings in Padang and Padang Pariaman,
because its epicenter was about 60 km offshore from Padang(Fig. 2(a)). As the Padang earthquake was an intra-slab
earthquake at intermediate depth with a comparable
magnitude, the event did not generate a tsunami ofsignificance [4]. Due to this earthquake, 1117 people werereported killed, 1214 severely injured, 1688 slightly injured,and 3 were left missing in West Sumatra. The earthquake also
destroyed many houses, buildings and infrastructure (heavily
damaged houses numbered 114,797, with 67,198 moderately
damaged and 67,837 slightly damaged). In Padang, 5458
buildings sustained damage [5]. This event occurred at theend of the working day, just 15 minutes after offices andschools closed; if it had struck earlier, the number of
causalities would definitely have been higher as a result ofbuilding collapses. Several hours after Padang earthquake, 1st
October 2009, Sumatran fault line generated Mw7.1 and
10km depth. Due to this earthquake destroyed many houses
and building (heavily damaged houses numbered 600, with
550 moderately damaged). Fig.2(a) shows Padang
earthquake and Kurinci earthquake.
There are four accelerometers in Padang. Three were donatedby Engineers Without Borders Japan (EWBJ) and installed in
2008, and the other was installed by the Indonesian
Government’s Bureau of Meteorology, Climatology andGeophysics (BMKG). However, only one ground motion
record is available for the Padang earthquake. Due to anelectric power cut during the earthquake, only the BMKG
device recorded the time history of the earthquake. The

observed record shows about 20 s of strong shaking with apeak ground acceleration (PGA) of 0.3 g and a predominantperiod of 0.5 s (Fig. 2(b)). response spectra at law period isgreater then Indonesia code for rock condition (0.83g)(Fig.2b).The location of this station is a mountainous suburb
about 12 km in from the coast. The subsurface condition at
this station is rocky; the average shear wave velocity for theupper 30 m of the subsurface here, Vs30, is 1200 m/s [6].

LegendPadang Mw7.6
Kurinci Mw7.1

PadangKurinci

(a)
(a)

(b)

(c)
Fig.2 Seismicity and time series. (a) two giant earthquake in
Padang and Kurunci, (b) time series of Padang earthquake, (c)response spectra due to Padang earthquake.

426

The paucity of measured ground motion information for thisearthquake results in a need for transforming observed data(seismic intensity) into parameters that are more useful forengineering purpose (e.g., engineering ground-motion
measures)[7]. From seismic intensity, we can evaluate
historical earthquakes, assess seismic hazard and damage,correlate different intensity scales, and rapidly assess theseverity of ground shaking [8]. In additional, we performed a
12-site microtremor array observations to determine the shear
velocity of subsurface structures at several districts in Padang.
From these observations, we obtained relationship between
soil types, predominant period and seismic intensity.

3. DAMAGE FROM THE 2009 PADANG
EARTHQUAKEThe city of Padang covers an area of about 695 km2 and isdivided into 11 districts: B. T. Kabung, K. Tangah, Kuranji, L.
Begalung, L. Kilangan, Nanggalo, P. Barat, P. Selatan, P.
Timur, P. Utara, and Pauh. 51.0% of the land is forested,
28.52% is used for farming, 9.54% for housing and 7.1% forrice fields [9]. The population of more than 857,000 isincreasing by 2% per year. The K. Tangah district has thehighest population and most extensive area compared withthe other districts in the city.
The central business area of Padang is close to the coast andconsist of several districts: P. Barat, P. Utara, P. Selatan and P.
Timur, B.T. Kabung, K. Tangah. The downtown area isutilized as a center of political and commercial activities.
Although the Padang earthquake affected all districts of thecity, the major damage occurred downtown, because about
80% of population lives near the coast.
The majority of houses in the city are one- and two-storey
non-engineered structures. These structures are typically builtof confined masonry, with reinforced-concrete (RC) frames
acting as confinement for the brick masonry walls. There arethree general categories of houses in Padang: permanenthouses (RC), semi-permanent houses (mix of RC and wood)and traditional houses (wood). Unfortunately, no detailed
damage statistics are available for each type of building, so
we cannot classify the category of the house.
This earthquake also affected lifelines in Padang. The strong
ground shaking destroyed public water distribution pipes
leading to 2,906 reported leakage points in total [10]. Damage
to pipelines forced the cessation of water delivery toconsumers for several weeks.
4. SITE CHARACTERIZATION BY MICROTREMOROBSERVATION

correspond to the resonant period of the site. This method
postulates the shape of the Fourier spectrum [11].
Equation. (1) shows the method used to calculate HVSR
using the observed records.

HVSR

1

and
denote the Fourier amplitude ofwhere
the NS, EW and UD components of each interval,
respectively, and is the frequency.
We performed 110 single site surveys that sampled every
district of the city of Padang. These observations were carriedout in November 2008, September, November, and December
2009 and January 2010. The locations of observations areplotted in Fig.3. Microtremor was measured using a GPL6A3P sensor. The two horizontal (NS and EW) and thevertical (UD) components were recorded simultaneously for
10 minutes with a 100 Hz sampling frequency.
We estimated the distribution of the peak periods of the
HVSRs for all sites in Padang using the ordinary kriging
technique (Fig.3). From single observations, we obtained a
predominant period of 2.0 to 4.0 s in the central business
district and less than 1.0 s in the mountainous areas. These
results indicate an affect related to the thickness of alluviumin the coastal area of Padang city, which decreases inthicknessinland.

4.1 SINGLE OBSERVATIONS

Fig. 3 white circles are single observation sites, result HVsRS
ratio

A microtremor is a very small ground motion that can berecorded on the ground surface. It can be produced by avariety of excitations (e.g., wind, traffic, breaking sea waves).
A full microtremor record can be described by one verticaland two horizontal components. Our analysis was conductedusing the recorded microtremor. First, the horizontal andvertical spectrum ratios (HVSR) were computed for all sites(Fig. 3). The peak period of the HVSR is known to

4.2 MICROTREMOR ARRAY OBSERVATIONS
The velocity of surface waves is well known to vary as afunction of frequency (or period) due to dispersion. Since
dispersion is a function of subsurface structure, thesubstructure can be estimated from a Rayleigh wave
dispersion curve. We carried out microtremor array427

investigations using 12 sites at several districts in Padang(Fig.4(a)). Dispersion curves were calculated using the SPAC
method [12] to obtain a velocity structure from themicrotremor recordings. An outline of the procedure follows.
It is necessary to simultaneously record microtremors with an
instrument array of at least three stations. The dispersion of ameasured surface wave is a response to the subsurface
structure directly below the array, and the estimation of thesubsurface structure causing the dispersion is determined bymeans of inversion of Rayleigh waves. The basic principlesof the SPAC method assume that the complex wave motionsof microtremors are stochastic processes in time and space. A
spatial autocorrelation coefficient for a circular array can thenbe defined when the waves composing the microtremor (i.e.,the surface waves) are dispersive. Hence, the spatial
autocorrelation is a function of phase velocity and frequency.
Rayleigh wave records were measured for the 12-arrayobservation sites using the SPAC method and inversion
analysis was undertaken on the observed dispersion curves toestimate the soil profiles. In the inversion analysis, theParticle Swarm Optimization (PSO) algorithm was adopted tosolve the non-linear optimization problem [13]. The basic
procedures of PSO are outlined below.
We estimate the subsurface structure of the model by solving
a nonlinear minimization problem with the fitness function
below.
2

3

whereis particle velocity of thecomponent inis the particle position ofdimension d in the interaction,
thecomponent in dimension d in interaction, and areis the best position achieved byconstant weight factors,
particle ,is the best position found by the neighbor ofparticle , and are random factors in the [0,1] interval andis the inertia weight. Before performing the inversion
analysis, the subsurface structure was assumed to consist ofhorizontal layers of elastic and homogeneous media above asemi-infinite elastic body. The shear
wave velocity and thickness of each layer are the parameters
determined by the inversion analysis. The results enable us todetermine the condition of shallow subsurface structures [14].
The outline of the SPAC method for the phase velocity
calculation of Rayleigh waves follows. The spatial
autocorrelation function is defined as∅ , ,0,0, ,∗, , ,The spatial autocorrelation coefficient is defined as theaverage of the autocorrelation function in all directions
over the circular array:
1
∅ , ,
5
,
2 ∅ 0,
where ∅ 0,is the SPAC function at the center 0,0 of
the circular array. By integration of the Equation. 5 we obtain
6

,

where
is the zero-order Bessel function of the first kindof x, andis the phase velocity at frequency . The
SPAC coefficient,
can be obtained in the frequency
domain using the Fourier transform of the observed
microtremors.
,

1
2

, ,
, ,

.

7

and
, , are the power densities of thewheremicrotremor at sites C and X respectively, and, , isthe cross spectrum between ground motions at these two sites.
Thus the SPAC coefficients may be obtained by averaging thenormalized coherence function defined as the spectrum
between points C and X in the direction . From the SPAC
coefficient, , the phase velocity is calculated for every
frequency from the Bessel function argument of Equation. 6,
and the velocity model can be inverted.

4

where
is the average velocity of the wave in the time
domain, and the harmonic waves of frequencyof themicrotremor have the velocity wave forms 0,0, , and, , , , observed at the center of the array 0,0 at
point, on the array.

428

Altitude(a)5.2 Estimation of JMA Intensity
The calculation of the seismic intensity determined by
the questionnaire is described as:
∑

IQ

8

where
is chosen by a respondent for the th question (e.g.,i=12 and=12),is theis a seismic coefficient,
number of effective questions of each questionnaire andisthe number of responses from all questions. To obtain the
JMA intensity (, the result from Equation. 8 is inserted
into Equation.
.

.

9

.

Finally, to obtain the JMA intensity
for each district,the calculation for each questionnaire is averaged as follows.
1(b)
Fig.4 Observation sites and distribution of average shear
wave velocity, (a) Array observation sites, (b)Contour of Vs30 (m/sec)

5 SHAKINGCHARACTERISTICS

10

where N is number of questionnaires in one location, andis the seismic intensity for each questionnaire.
BY

QUESTIONNAIRE SURVEY
5.1 Survey Outline
Seismic intensity has been used to quantify the severityof the ground shaking based on the observed or felt effect in alimited area [7]. The method was originally developed by[15] and has been widely applied to the development ofseismic macro or micro zoning maps in Japan since 1970. The
method has also been applied in India and Indonesia by
Honda in 2005. It has been useful for estimating seismic
shaking in a limited area and for determining intensity
distributions over local areas [7]. The original questionnaire
sheet was written in Japanese; however, it was translated into
Indonesian, for its application to Aceh. The questionnaire has
35 items that cover recognition of shaking, location, shaking
duration, possibility of movement, structural damage,swinging of hanging objects, and 27 other items.
For the 2009 Padang earthquake, some sentences weremodified to make it more relevant for local people while notchanging the original topics covered by the questionnaire.
People living near the observation points were interviewed
using the questionnaire [15]. The questionnaire survey was
conducted from December 24 to 31, 2010, three months after
the main shock. The survey was carried out in all districts ofPadang by distributing and completing 500 questionnaires
through a direct interview process with residents of the city.
The interviewers explained each item of the questionnaire toresidents, discussed the responses given, and documented theanswers on the standard answer sheets.

The results of the questionnaire survey conducted to estimate
the shaking intensity distribution in Padang during theearthquake are summarized in mapped. The results of thequestionnaire survey conducted to estimate the shaking
intensity distribution in Padang during the earthquake are inmapped. The seismic intensity (IJMA) for the suburbs anddowntown were lower 5 (5-) and upper 5 (5+) respectively(Fig.5). A JMA seismic intensity, IJMA, value of 5corresponds to 4.5< IJMA <5.0 and 5+ equates to 5.0< IJMA <5.5.
A value of 5+ corresponds to very strong ground motion
where many people are considerably frightened and find it
difficult to move. Non-engineered structures sometimes
collapse, walls crack, and gravestones and stone lanterns
overturn. A value of 5- corresponds to strong ground motion.
Many people are frightened and feel the need to hold on tosomething stable. Occasionally, less earthquake-resistantbuildings suffer damage to walls, and windows may breakand fall. The distribution of various intensity values camefrom differences in the subsurface structural conditions ofeach district.
5.3 SHAKING CHARACTERISTICS
By comparing the results of the 500 questionnaires of theseismic intensity survey with microtremor observations (110single-site observations and numerous other 12-site arrayobservations), we found a correlation, in which thesubsurface area or site is seen to correspond to the soft soil
condition (which we here define as Vs30<400 m/s) thatexceeds the predominant period. This area corresponds to thethick alluvium in the coastal area of Padang.

429

the coastal area to rocky conditions in the mountains. The
seismic intensity decreases from the coastal area (5+) tomountains ( 4+).
These results provide critical information for making shaking
maps, updating hazard maps, and developing disasterprevention countermeasures in Padang.
REFERENCES

Fig.5 Average seismic intensity
The thickness of the alluvium gradually decreases in alandward direction from the coast. As a result, greater seismic
intensities are observed in the coastal area and the value
decreases from coastal to mountain areas. Important
information from respondents’ answers were collated. The
area was divided in two by considering subsurface structure
condition: shear velocities Vs30<400 m/s, corresponding tosoft soil (area 1), and Vs30>400 m/s, corresponding to theengineering bedrock conditions (area 2).
The structural type, the age of building, and the number offloors in each of the two areas were found to be generally thesame. However, the shaking duration was significantly
different.
In area 1, 96% of respondents said that the duration of
shaking was longer than 2minutes, but in area 2, only 39% of
respondents felt the duration of shaking to be longer than 2
minutes. These observations support the existence of a thickalluvial layer that caused the prolonged shaking that mostpeople living in the coastal area felt during the earthquake.
6 CONCLUSIONS
Our survey conducted in Padang consisted of (1) microtremor
observations (single and 12-channel arrays) made before andafter the earthquake of September 30, 2009, and (2) aquestionnaire survey of 500 people in the Padang area. The
central part of the city, consisting of the four districts, P.
Utara, P. Barat, P. Selatan and P. Timur, experienced greater
seismic intensities (5+) compared with other areas (districts)
of Padang.
According to microtremor observations, downtown Padang isunderlain by soft soil conditions (Vs30<400 m/s). Consistent
results concerning the soil condition were found based on
predominant period observations and the questionnaire
survey. In both cases, the coastal area was determined to havea soft soil conditions (Vs30<400 m/s), a longer predominantperiod, and a greater seismic intensity.
Padang has a thick alluvial layer in the coastal area (with apredominant period between 2.0 and 4.2 s) that thins toward
the mountains (with a predominant period less than 2.0 s).
The subsurface geology also changes slowly from soft soil in[1] Prawirodirjo, L., Y. Bock, J.F. 2000, One Century ofTectonic Deformation along the Sumatran Fault from
Triangulation and Global Positioning System Surveys, J.
of Geophysical research, 105, 28, 343-28,363.
[3] Natawidjaja and Wahyu Triyoso 2007. The Sumatran
Fault Zone-from Source to Hazard, J. of Earthquake andTsunami, Vol. 1 No. 1, 21-47.
[4] EERI 2009. The Mw 7.6 Western Sumatra Earthquake of
September 30, 2009, Special report.
[5] BNPB 2009. Total Damage Report and Verification forWest Sumatra due to Padang Earthquake, BNPB report,
2009[6] Rusnardi, J. Kiyono, Y. Ono. 2010, Seismic Hazard
Analysis for Indonesia, Proc. of International
Symposium on a Robust and Resilient Society against
Natural Hazards and Environmental Disasters and thethird AUN/SEED-Net Regional Conference on
Geodisaster Mitigation, pp.317-325.
[7] Tselentis. G-Akis and LaurentiuDanciu. 2008. Empirical
Relationship between Modified Mercalli Intensity andEngineering Ground-motion Parameters in Greece, bull.
Of the Seismological Society of America, vol.98, No.4,
pp.1863-1875[8] Wald, D.J., V. Quintoriano, T.H. Heaton, H. Kanamori,
C.W.
Scrivner,
andC.B
Worden(1999).
TriNet”ShakeMaps”: Rapid Generation of Peak Ground
Motion and Intensity Maps for Earthquake in southern
California, Earthquake Spectra 15, no. 3, 537-555.
[9]
Pemerintah
kota
Padang.
availableat
atwww.Padang.go.id(Padang local government
website).
[10] Padang spring water agency, www.pdampadang.co.id[12] Aki, K. 1957. Space and Time Spectra of Stationary
Stochastic Waves, with Special Reference toMicrotremor, Bull. Earth. Res. Inst., Vol. 35, No. 3,
415-456.
[13] Keneddy, J. and Eberhart, R. C. (1995), Particle Swarm
Optimization, Proc. Of IEEE International conference on
Neural Networks, Vol.4,pp.1942-1948.
[14] Ono, Y., Kiyono, J., Rusnardi, P. R. and Noguchi, T.
2010. Microtremor Observation in Padang City,
Indonesia to Estimate Site Amplification of Seismic
Ground Motion, Proc. of International Symposium on a
Robust and Resilient Society against Natural Hazards
and Environmental Disasters and the third
AUN/SEED-Net Regional Conference on Geodisaster
Mitigation, pp.386-391.
[15] Otha, Y., Goto, N., and Ohashi, H., (1979). A
questionnaire Survey for Estimating Seismic Intensity,
full. Fac. Eng., Hokkaido University 92, 241-252 [inJapanese]

430

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Groundwater Contamination Due to Irrigation of Treated Sewage
Effluent in the Werribee Delta
Hiroyuki II, Masanobu TANIGUCHI, Ataru SATOH, Faculty of Systems Engineering, Wakayama
University, Japan
Graeme ALLINSON, Matt KITCHING, George CROATTO, Bruce SHELLEY, Department of Primary
Industry, Victoria, Australia


ABSTRACT: The Werribee River, groundwater and Melbourne’s
treated sewage effluent (or recycled water) are used for irrigationwater in the Werribee Delta. Groundwater beneath the Werribee
delta may be contaminated by the recycled water. The mixing ratiosof the sea, river and recycled waters in groundwater were calculated
from water chemistry and isotopic ratio. Inland of the Werribee
Delta the river mixing ratio was very high (70 %) and sea andrecycled water mixing ratios were low. At the delta area, someborehole water had a high river mixing ratio, and others a low river
mixing ratio and high recycled water ratio (over 60 %). Overall, atthe delta area, half of the groundwater had a high mixing ratio ofrecycled water. The marine origin for Cl- in the groundwater wasvery limited, and was only found at the seaward boundary of thedelta area. Origin of NO3- for groundwater was recycled water andfertilizer at the delta.

Keywords: Oxygen isotope, Recycled water, Sewage
irrigation, Estimation of groundwater origin
1. INTRODUCTION
1.1 Study area and Purpose
The Werribee delta area at the west of Melbourne, Victoria,
Australia is important for growing vegetables for the city of
Melbourne. However, because of small amount ofprecipitation in the area (average 500 mm per year [1]), theWerribee River, groundwater and recently treated sewage
waters from Melbourne (recycled water) have been used for
irrigation water. The groundwater used in the delta area ispumped up from 6 to 30 m in depth. A decline ofapproximately 2m in the aquifer level, and increasing salinity(varied from 2500 to 9000 EC) led to bans and restrictions on
ground water use in 2003, in part to mitigate saline intrusion[1]. In 2004 the use of recycled water from Melbourne’s
Western Treatment Plant began. Approximately 90 % of theirrigators had signed on to receive recycled water in 2008.
Although average salinity of recycled water was 2000 EC andless than groundwater EC, groundwater contamination
caused by irrigation water is a concern because the recycledwater contains trace elements and high nitrogenconcentrations. Therefore, in this study, origin of thegroundwater was estimated using water chemistry including
isotopes and influence of recycled water on groundwater was
clarified.
1.2 Recharge area
The upper catchment of the Werribee area consists of Silurian
sandstone and mudstone, Devonian granite, Permian glacial

tillite, older Tertiary basalt and recent alluvial sand; themiddle stream area consists of Pliocene basalt; and the lower
stream area consists of recent alluvial sediments [2]. The
basement of the middle to lower stream area is Tertiary
sediment covered with basalt lava and recent alluvial
sediments [3]. The Tertiary sediments are composed of sand,clay and gravel including older volcanics and have high water
permeability. The recharge area of delta groundwater was
thought to be from the mid-catchment to delta area.
2. METHOD
Sea, river, borehole, recycled and rain waters were studied
from 2007 to 2010. Sea water was sampled at the delta coast.
Borehole waters were sampled from the middle stream area tothe delta area. The Werribee River and its surrounding riverwaters were sampled. Recycled waters were sampled from
irrigation supply channels and drains in the delta. Rain water
was sampled at Werribee. Temperature, pH ORP and EC
values were measured on site using a portable instrument(HORIBA, pH meter D-54, 55). Samples for anion and cation
analysis were filtered through 0.45μm filters and were
measured using liquid ion chromatography (DIONEX,DX-ICS, AQ-1500 and ICS-1600). Also HCO3- wasmeasured by an acidic titration. Trace elements in the samples
were analyzed by ICP-MS (Ultramass ICP-MS, Varian
Australia). The oxygen and hydrogen stable isotope ratios ofwater samples were measured using a mass spectrometricsystem (Finnigan Mat Delta Plus), followed by the CO2-H2Oand H2-H2O equilibration technique using platinum catalyst.
The measured isotopic ratio was calculated using equation(1).

  ( R x / Rst  1)  1000

(1)where Rx is the stable isotopic ratio of x，and Rst the stable
isotopic ratio of st
3 RESULTS
Groundwater sampled from the boreholes at the delta isthought to be derived from the recycled water, the sea andupper stream recharge water. To estimate the origin of delta
groundwater, sea, recycled, and river waters were analyzed.
Geologically, the upper Werribee River catchment consists ofhard impervious rock, and the mid–catchment is thought to be
the recharge area for delta groundwater. River water in the
mid-catchment was thus regarded as the recharge water fromthe upper catchment in data analysis.

431

200

NO3- (mg/L)

150

50

0

(2)

sea

recycledriver

Fig.1 NO3- concentration for the sea, recycled and riverwaters
2.5

2

Ｆ- (mg/L)

1.5

1

0.5

0

sea

recycledriver-

Fig.2 F concentration for the sea, recycled and river waters
25000

20000

3.2 Estimation method of mixing ratio
First, because the Cl- concentration of the sea was extremely
high relative to other waters, Cl- was used for determining themixing ratio of sea water. The mixing ratio between sea andriver water (recharged water) was calculated using equation(2).
Msea1 = (Cl-sam-50)/(19000-50)*100%

100

Cl- (mg/L)

3.1 Character of water chemistry for sea, recycled waterand rivers
Fig.1, 2 and 3 show the NO3-, F-, and Cl- concentrations of sea
water sampled in Port Philip Bay, recycled water from deltadrains and channels, and river water sampled at the Werribbe
River, Little River, Skelton River, Lerderderg River, Pykes
Creek, and Toolern Creek (excluding the lower stream area).
The recycled water has extremely high concentrations of
NO3- (over 100 mg/l). The mixing ratio changes withsampling site and sampling time and so NO3- concentration isvariable. A high NO3- concentration value (100 mg/l) was
adopted as the recycled water concentration for data analysis.
As the NO3- concentration of the river and sea waters, andmost groundwater at the delta were very low, the main sourceof NO3- is thought to be treated sewage effluent. A low NO3concentration value (0 mg/l) was adopted as the sea and riverwater NO3- concentration for data analysis. So, NO3- was agood indicator of mixing ratio of recycled water because sea
and river water concentration were 0 mg/l. However, NO3- forgroundwater is contaminated by fertilizer at a farm land andcontamination must be considered for determining mixing
ratio using NO3-.
F- in the recycled water was much higher than in sea and riverwaters. The origin of F- for the recycled water was thought tobe derived from sewage and F- is stable. So, F- was also agood indicator of mixing ratio of recycled water. However, as
F- concentrations for recycled and river water were variable,
F- was not used for data analysis.
The sea water had an extremely high concentration of Cl(19000 mg/l), relative to the recycled water (450 mg/l), andthe river water (50 mg/l). Generally, Cl- was stable and thenCl- was a good indicator of mixing ratio of sea water.
Fig.4 shows that Oxygen isotopes for the Werribee River andits surrounding river waters, channel water (recycled water
before irrigation), and drain water (after irrigation). From thefigure, oxygen isotopes were -3.3 per mil for the recycledwater and -5.8 per mil for the middle catchment river water.
Oxygen isotopes were generally 0 per mil for the sea water.
-5.8 per mil for average oxygen isotope of rain water at thedelta area was in agreement with those for the middle
catchment river water and then oxygen isotope for thecatchment area was thought to be -5.8 per mil. Since theisotopic ratio of oxygen for each type of water was different
and oxygen isotopic ratio was stable when water reacts withminerals at the low temperature condition, the oxygenisotopic ratio was a good indicator to determine mixing ratios.

15000

10000

5000

0

sea

recycledriver

Fig.3 Cl- concentration for the sea, recycled and river waters

432

4
100
90

2

80

mixing ratio (%)

delta 18O (permil)

70

0

-2

-4

60

sea
recycled
river

50
40
30
20
10
0

-6

werribee 2007/12/1
werribee 2008/4/1
werribee 2008/9/1
werribee 2009/7/1
werribee 2010/1/1
Other Rivers
Drain at the delta
channel water

-8

-10
0

10

20

30

40

0

2

50

60

6

8

10

12

14

100

70

Distance from coast (km)80

mixing ratio (%)

Fig.4 Delta 18O for recycled and river waters
where Cl-sam is the Cl- concentration of borehole water. The
mixing ratio between the sea and recycled water was
calculated using equation (3).
Msea2 = (Cl-sam-450)/(19000-450)*100%

60

sea
recycled
river

40

20

0

(3)

0

5

10

15

20

distance from coast (km)The actual value is between the Msea1 and the Msea2. The
difference between the Msea1 and the Msea2 was very small
and the average value was adopted.
Nitrate (NO3-) is also an important indicator because the sea
and river concentrations are very low relative to the recycledwater, and NO3- concentration is very sensitive to the mixing
ratio of the recycled water. The mixing ratio between therecycled water and river water (recharge water) was
calculated using equation (4).
Mrecy= NO3-sam/100*100%

4distance from coast(km)100

mixing ratio (%)

80

60

sea
recycled
river

40

20

(4)

0

where NO3-sam is NO3- concentration of borehole water. If
Mrecy was less than 100 %, the Mrecy value was adopted as a
mixing ratio of the recycled water. However, when the Mrecy
value was over 100 %, the Mrecy value cannot be used. Inthat case, the groundwater is thought to be contaminated bynitrogen fertilizer from farm land on the delta. When both the
Msea and Mrecy values can be derived, all mixing ratio of sea,recycled, and river waters can be determined. The accuracy ofthis process was confirmed by isotope values. The difference
between the calculated groundwater oxygen isotopic ratio
from the determined mixing ratio of sea (0 per mil), recycledwater (-3.3 per mil), and river water (-5.8 per mil), and themeasured isotopic ratio was compared. When the difference
between both the isotopic ratios was less than 2 per mil, themixing ratio was fixed. When the Mrecy is over 100 % or thedifference is over 2 per mil, mixing ratio cannot be
determined and the limited values only were used.
3.3 Estimation results of mixing ratio
Fig.5 shows the mixing ratios of the sea, river and recycledwaters for borehole water in December 2007, April 2008, July

0

10

20

30

40

Distance from coast (km)

100
90
80

Mixing ratio (%)

70
60

sea
riversrecycled

50
40
30
20
10
0
0

10

20

30

40

Distance from coast (km)Fig.5 Mixing ratios of the sea, river and recycled waters forborehole water in December 2007, April 2008, July 2009 andJanuary 2010.

433

2009 and January 2010. The river mixing ratio increased withthe distance from the coast. Inland, where the distance fromthe coast is over 10 km, the river mixing ratio was very high
(70 %), and the sea and recycled water mixing ratios were low.
In the delta area, where the distance from the coast is less than
10 km, there were two types of borehole waters. One had highriver mixing ratio and the other low river mixing ratio. In the
low river mixing case, the recycled water ratio was high (over60%), suggesting direct surface water intrusion. Someborehole values indicated high mixing ratio of the sea waterin the delta area. Overall, in the delta area, half ofgroundwater has high mixing ratio of the recycled water.
On the other hand, inland area over 10km distance from thecoast, the analyzed mixing ratio of sea and recycled water
were very low. Inland area, recycled water was not used andfar from the sea. Therefore, the analyzed results were inagreement with the actual phenomenon.
Fig.6 shows the mixing ratios of sea, river and recycledwaters for borehole water with depth at the delta. The mixing
ratio was not clearly divided by depth. The recycled water
ratio was always high at the delta, although some borehole
water has low mixing ratio of the recycled water. The sea
mixing ratio increased with depth.

some borehole water was thought to be the recycled water.
However, on the calculation of the Mrecy, the Mrecy value
for some borehole was over 100 % and so some borehole
water is thought to be contaminated by nitrogen fertilizer
from farm land on the delta.

3.4 Cl- and NO3- origin for borehole water
Fig.7 shows Cl- and NO3- concentrations with distance fromthe coast for borehole water. Although Cl- concentration
increased towards the coast in Fig.7, the sea water mixing
ratio did not increase close to the coast. The increase of Cl- in
the borehole water was not thought to be caused by salt ofmarine origin. NO3- concentration increased near the coast(Fig 7) and the recycled water mixing ratio increased towards
the coast. As the recycled water was only used for irrigationin the delta area and some borehole waters at the delta area
had high NO3- concentration, initially the origin of NO3- in

Fig.7 Cl- and NO3- concentrations of borehole water with
distance from coast.

1.5

5000
Ni(ug/l)Mn(ug/l)

1.2

4000

0.9

3000

0.6

2000

0.3

1000

metal(ug/l)

F(mg/l)F(mg/l)

0
0

10
20
30
40
distance from coast (km)

15000

0
50

500
Cl(mg/l)NO3-(mg/l)

400

9000

300

6000

200

3000

100

NO3-(mg/l)

Cl(mg/l)

12000

100

0
0

mixing ratio (%)

80

60

Fig.8 Relationship between F- andconcentrations in the borehole water.

sea
river
recyled

40

20

0
0

10

20

30
depth(m)

40

50

60

Fig.6 Mixing ratios of sea, river and recycled waters forborehole water with depth at the delta.
434

0
50

10
20
30
40
distance from coast (km)traceelement

In the mid-catchment area, the low mixing ratio of therecycled water was in agreement with low concentration of
NO3- concentration for borehole water. Initially, the source ofNO3- in the borehole water was thought to be the recycledwater. However, as the recycled water is not used in the
mid-catchment area, the NO3- concentration for rivers wasvery low, and some high NO3- concentration borehole water
was found in that area, the high NO3- concentration is thought
to be caused by nitrogen fertilizer [4] .

3.5 Origin of F- and metal
Fig.8 shows the relationship between F- and trace element
concentrations of the borehole water. The F- concentration forthe borehole water increased near the coast and reached 1.5
mg/l at the delta area. The F- concentrations for the sea andriver waters were 0.5 mg/l and less than 0.3 mg/l, respectively.
On the other hand, F- concentration for the recycled water washigh (1 to 2.5 mg/l). Therefore, mixing of recycled water was
thought to be important for high F- concentration. However,
upstream of the delta area, F- concentration for the borehole
water was also above 0.5 mg/l; in this case the increase of Fwas caused by contact with sediments. Overall, F- origin of
the borehole water was thought to be sewage or sediments.
At the delta area, borehole waters with high Ni and Mnconcentrations were found. Fig.9 shows Ni and Mnconcentrations for the sea, recycled and river waters. Bothmetals were low for recycled and sea water and high for theriver waters. The high Ni and Mn concentrations in borehole
water are thought to derive from river water which infiltrated
near the delta area (because river water recharge ratio washigh at the delta and surrounding area).
560

4 CONCLUSION
The mixing ratios of the sea, river and recycled waters inborehole water were calculated from water chemistry andisotopic ratios. Inland, the river mixing ratio was very high
(70 %), and the sea and recycled water mixing ratios were low.
At the delta area, some borehole water had a high river
mixing ratio, and others low river mixing ratio with highrecycled water ratio (over 60 %). Overall, at the delta area,
half of groundwater has high mixing ratio of the recycledwater.
The origin of each analyte for borehole water was estimatedby considering the mixing ratios of the samples. The marine
origin for Cl- in borehole water was very limited and was only
found in the delta area. The origin of NO3- for borehole waterat the delta is thought to be recycled water and/or fertilizer. In
the mid-catchment area, fertilizer was the only NO3- source.
The origin of F- in borehole water is thought to be recycledwater in the delta area, and sediments in the mid catchment.
The high Ni and Mn concentration in borehole water isthought to derive from river water which has infiltrated into
the aquifer near the delta area.
5 REFFRENCE
[1]
[2]
[3]
[4]

140
Ni (ug/L)
105

Mn (ug/L)

280

70

140

35

0

Ni (ug/L)

Mn (ug/L)

420

0
sea

recycled

riversFig.9 Ni and Mn concentrations for sea, recycled and riverwaters.

435

Department of Primary Industry, Victoria, “Atlas of Western Irrigation
Futures”, Southern Rural Water, Victoria, Australia, 2009. pp1-36.
Condon MA, “The Geology of the Lower Werribee River, Victoria”
Soil Mechics, 4th ed. vol. 63, Proc. Royal Society Victoria, Australia,
1951, pp. 1–25.
John L, “Port Philiip Region Groundwater Resources- Future Use andManagement” Department of water Resources Voctoria, Australia.
1992. pp1-116.
Commonwealth Scientific and Industrial Research Organisation,“Groundwater Nutrient and Toxicant Inputs to Port Phillip Bay”,
Technical Report No.13, Victoria, Australia, 1993. pp1-63.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Process of Invading the Alien Plant Species into the River Ecosystem
Michiko Masuda1, Yumie Ito1, Fumitake Nishimura.2, 1Department of Civil Engineering, Nagoya Institute ofTechnology, 2Department of Environmental Engineering, Kyoto University.



ABSTRACT: A lot of alien plant species invaded the fluvial
environment. Up to now, some researchers pointed out that the
disturbance by the flood help the invasion of alien plants. But there
are a few studies how alien plants invade into the river ecosystem.
Because grand-scale experiment was difficult, considering the flood
control and water utilization of the river, the alien invasion had notbeen examined. Fortunately, we have a chance to experiment how
the alien species invaded by using the park where the buried seed
was removed. At the construction of the park, Sicyos angulatus must
be removed from this area. This species make a lot of burial seedpopulation, then the planner of the park removed soils 2m in depth
and brought into clean soils from deeper area. As a result three of thefollowing became clear, 1) the vegetation of the river wasmaintained by seeds flow out, 2) the seeds were supplied with risingof the river, 3) the duration of the seed emerging at alien plant
species are longer than that at native plant species. The character thatthe seeds are able to emerge anytime and anywhere like the alienplants species are advantageous at the fluvial environment whererising of the river sometimes occurs. It was shown that the seed
emergence occurred in a short term like native plant species missed
the occupation chance. This pattern was remarkably admitted inspring in autumn.
Keywords: Seed emergence, Invading plant, Germination
characteristics, Native plant.
1. INTRODUCTION
The river law of Japan was revised in 1997. A past river law
was aiming at the flood control and the water utilization.
However, environmental preservation was added to the
purpose in 1997. Law revision is received, and "Multi-naturalriver-making" has been done as a stream management thatconsiders the environment. "Multi-natural river-making"means the river is managed for maintenance and the creationof a native habitat of the living thing that the river originally
maintains and various river spectacles. The protection of thehabitat of wildlife and the approach of the aquaticconservation are performed on the many site of"Multi-naturalriver-making".
Many“Multi-naturalriver-makings” constructions are promoted after 1997, andrestoration of natural condition is expected [1].
In “Multi-natural river-making", first of all, it is necessary
to recover vegetation. Vegetation is recovered because of
making the habitat of other living things, and leading to therecovery of the ecosystem. Three methods are given to thevegetative restoration [2]. First, it is a method of leaving to
the seed scatter of around natural vegetation. Second, it is amethod of transplanting and sowing suitable plant species.
Third, it is a method of utilizing seed bank around there. If an
appropriate seed source to surroundings doesn't exist, the first
technique cannot be used. In the second technique, it is aproblem that the selection of an appropriate plant is difficult.
The third method is being paid attention to as a method ofmaking use of the potential, natural recovery power in theregion [3]. However, the fluvial environment is a hard
disturbance area, and the transportation power of earth andsand is also large. Therefore, the examination which
technique was suitable had not been done.
Moreover, another big problem exists for the vegetative
restoration of the fluvial environment. It is a disturbance of
biodiversity by the invasion of the alien plants. Rising forms
the bare land in the river area. The bare land is formed
frequently. The bare land can be invaded at any time by theplant with a life history of ruderal species. The invasion of analien plants have been diminishing the biodiversity at theriver condition. Especially, the alien plant that invades thefluvial environment brought in to Japan by the seed medium.
Therefore, genetic diversity is very high which adjusts easily
to the each invading environment [4]. Moreover, many alienplants brought in exotic area without their disease or natural
enemy at all. Therefore, the activities of them tend to be highmore than native species [5].
The Ministry of Land, Infrastructure and Transport that hadsuch a problem founded the vocational training center named“The School in the Yada River Waterside” using“Multi-natural river-making” in Yada river at central Japan. It
gropes for shape that the biodiversity can be maintained when
the diversity of the river is created in this facilities, andvegetation restoration is done. Then there were seeds of Sicyos
angulatus, the species notorious alien species, in this area. So
to recover the rich biodiversity of the river, all the surface soil
with burial seeds was buried under the depth of 2m, and theburied soil under the 2m depth of there with no seed bank wascovered over there.
Then, it was investigated how the vegetation restoration of
the river was done in the “Multi-natural river-making”. 1)
Where is the seed supplied when the buried seed doesn't
exist? 2) When do the seeds of many kinds of species invade
into the bared area? 3) Is there a difference in the method of
the invasion in the alien and native species? It was paid
attention to these three points, and analyzed.

2. MATERIALS AND METHODS
2.1 Study site
The study was carried out on a left bank of the Yada Riverat Joganji, Nagoya City (35o 12’ 58’’N, 136o 55’ 2’’E,
altitude 9 m). Yada river is a urban river in the Shonai River
water system of the extension 23km and 115 km2 in the valley
area. The putting substitution construction of the river
channel was done in 1932. Today’s Yada River is a circle belt

437

placed between Shonai river and Yada River until the
beginning of the Showa era. Because this region had
frequently received the flood damage, Yada River put up to apresent river channel and was substituted by repair work.
At the study area, the biotope creation business was done in2008. This biotope is named “The School in the Yada River
Waterside”. It is a place of the activity of children's
environmental study and nature experiences now.
2.2 Vegetation area analysis
The region was made land with a cleared surface as a businessin 2007, and the buried seed was removed from this area.
Then, it took 8 pictures of the entire study area from April to
December in 2007. We divided the study site 200 meshes (2mx 2m) using the pictures. The class of each vegetation rate offour ranks was installed, and we analyzed it at the time ofpassing. The rank of the vegetation rate ranging 0~9% isassumed as A, and 10~89% is assumed as B and 90~100% isassumed as C. The part of starting constructing was assumed
D rank, and excluded from the analysis. We divided eachmesh into the three categories depending on a place, theoutside, the inside, and the waterside, and analyzed it at theinvasion season using Tukey-Kramer test.
2.3 Seedlings emergence census
The biotope was made in the investigation place at February
in 2008, where all the buried seeds were removed from thesoil within 2m in depth. For seedling emergence census, 10
quadrats (80 x 50 cm) were set at every 5 m in waterside part,low water bed part, and high water bed part on thehydrophilicity shore protection slope. The total area was
12m2. The species of emerging seedlings was identified bycomparison with the seedling specimen which had been
prepared by germinating the collected seed under laboratory
conditions. The recorded seedlings were checked by pigmentink. Census was carried out from April 5’2008 to May
20’2009 at 2 weeks intervals except for the summer andwinter season when no emergence was observed in the
preliminary inspection. Unidentifiable seedlings constituted
only less than 1% of the total seedlings recorded during thestudy period except grass family. Using this data, it wascompared the duration of seedling emergence between alienand native plants by Mann-Whiteney test.
3. RESULTS
3.1 Spatial patterns of seedling emergence
In April 2007, there was no vegetation area in the study site.
The invasion of the plant body started gradually in May, andthe vegetation rate of all the meshes exceeded 10% in July(Fig. 1). Vegetation area was increasing from May to August,and was decreasing from September to November.
In order to clarify the invasion part of the bared area by theseedling, the study area was divided study site into three
sections, I is water site part (upper edge in Fig. 1), II is inside
part (all the mesh except outline mesh in Fig. 1), III is outside
part (edge mesh except upper edge in Fig. 1). It wascompared the time that each part reached B rank by
Tukey-Kramer test. There were some significant differencesamong three parts. I part was occupied by invading plants
earliest (t = -7.55, P < 0.001 compared II, t = -5.05, P < 0.001
compared III). And II part was invaded by plants later than III
part (t = -3.08, P < 0.01). It was also compared the time thateach part reached C rank by Tukey-Kramer test. There was no
significant difference between I part and III part. But II partwas the latest area occupied by invading plants among allparts. There were significant differences between II part and I,
III parts (t = -3.08, P < 0.01 compared I, t = -5.14, P < 0.001
compared III).
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Fig. 1. Seasonal patterns of invade bared area by seedlingemergence. Each rank showed vegetation rate, A rank is 0~9%
A, B rank is 10~89% B, C rank is 90~100% C. Upper side iswater area. D rank is construction part D

3.2 Temporal patterns for seedling emergence
The temporal patterns for each species were shown in Fig. 2-4.
Fig. 2 showed the patterns of high water bed part seedlingemergence that over three seedlings were observed. Here it
was observed the significant difference of seedlingemergence duration in spring between alien and native plants(U=22, N=21, P<0.05 in spring). But there is no significant
difference of seedling emergence duration in autumn between
alien and native plants (U=28.5, N=16, P=0.87). Fig. 3
showed the patterns of low water bed part seedling emergencethat over three seedlings were observed in the census quadrats.
Here also it was observed the significant difference of
seedling emergence duration in438

Fig. 2. Seasonal patterns of seedling emergence from April
2008 to May 2009 at high water bed part. Shaded histogram
showed alien plants, open histogram showed native plant

spring between alien and native plants (U=8, N=15, P<0.05).
And also, there was no difference in autumn in low waterside
bed (U=10, N=12, P=0.305). Fig. 4 showed the patterns of
seedling emergence that emerged four seedlings at waterside
part in census quadrats. Most species emerged in spring or
autumn. There was no difference of seedling emergence
between alien and native plants (U =36, N=22, P=0.103 inspring, U=8.5, N=11, P=0.274).
In high water bed part, 35 alien species were observed in 10
quadrats, and total seedling number was 2232. And 35 nativeplant species were observed and total seedling number was
4753 (90% of the seedlings emerged in autumn). Total native
seedling number is superior to those of alien plants. And thevegetation survey was done where more than 50% vegetation
cover was some alien plants. This shows that the establishedrate of the seedling of the alien plant is higher than that of anative plant. Though it wasfrom a month to 2 months that emergence of the native plant
species in spring, alien plants seedling emerged in the studysite during 2 to 4 months. Especially seedlings of Erigeron
species were emerged from April 2008 to May 2009 except
August 2008.
In low water bed part, seedlings of 39 alien plants species
and 35 native plants species were observed. And the numberof seedlings was almost same, total seedling number of alienplant species was 2891 and that of native plants is 2614.
Almost of observed seedlings of native plants emerged inautumn. Compared with native plants, the seedlings of alienplants emerged both in spring and autumn. In the low water
bed part, seedlings of Trifolium repens emerged very longtime from April 2008 to April 2009 except August and
September. Compared with alien plants, native plants
emergence occurred in spring or in autumn. The short time
emergence patterns of native species were observed. More439

Fig. 3. Seasonal patterns of seedling emergence from April
2008 to May 2009 at low water bed part. Shaded histogram
showed alien plants, open histogram showed native plantseedling emergence.

than 80% of the vegetation cover was composed by alienplants in spring 2009.
In waterside part, the seedlings of 36 alien plants species
and 49 native plants species were observed. The total seedlingnumbers were 4333 in alien and 3458 in native plants. In thispart, the more than 50% vegetation cover was composed ofnative plants. The kinds of alien plants species is lower thanthat of native plants species compared with the other parts.
Alien plants emerged through a year, especially T. repens andErigeron species, there was not serious invasion by alienplants observed.
4. DISCUSSION
Recently, the weed risk assessment (WRA) is paid attention
as a risk evaluation system of the species before alien species
intentionally introduces into Japan. In Australia the operation
results of the WRA system have exceeded ten years [6].
Because the damage of the exoticism to biodiversity has
become the largest problem worldwide, the WRA begins to
be used also in Japan [7]. In the WRA, there is a check item
concerning the item concerning a character of ruderal strategy.
Actually, the invasion of alien species with ruderal strategy
become a big problem in Japan.

Especially the invasion of alien species in the river system
where continuous disturbance of each rainfall occurred is theserious problem [8].
From 1997 many biotopes have been constructed
everywhere in river by using “Multi-natural river-makings”
by the Ministry of Land, Infrastructure and Transport. In
every biotope there is the problem of alien species more or
less. Eragrostis curvula, Coleopsis lanceorata and S.
angulatus invade biotopes and cause a serious situation tomodify vegetation [9]. These plants were positively
introduced for preventing sand erosion and greening in theslope. The seeds amount of E. curvula was 30689 kg per year
from 1991 to 2000 in Japan (from the Ministry of Agriculture,
Forestry and Fisheries materials). The regeneration of these
species has succeeded and a huge amount of seeds have been
supplied by Rising water level. It is known that it depends
whether the alien species can invade on existing plantvarieties [10] [11].
The biotopes by using “Multi-natural river-makings” hadno vegetation cover when biotope constructions werecompleted. And there was no seed bank in the new biotopes,
because the necessity for preventing the alien species
invasion was requested in The Invasive Alien Species Act
from 2005. There are many burial seeds of invasive alien440

Fig. 4. Seasonal patterns of seedling emergence from April 2008
to May 2009 at waterside part. Shaded histogram showed alienplants, open histogram showed native plant seedling emergence.

plants species in every river. Then the construction must be
done without many burial seeds.
Where is the seeds carried when the construction of thebiotopes without burial seeds was done? A rising of riveralways transports the seeds. Fig. 5 shows the line of seeds
dispersal where the seeds of L. juncea var. subsessilis
emerged along the water level of the rising. Always seeds
have been transported by rainfall. But the intensity of plantinvasion was different according to the season. In Japan,
summer season is too hot when the survival rate of seedlings
is low. And also winter season is too cold when the seedlings
cannot survive for freezing [12]. Most emergences occurredin both spring and autumn. The seeds of native species
emerged very short duration in spring, and relative long
duration in autumn. Compared with native species, the alien
species emerged for very long time both in spring and
441

Fig. 5. Seedlings lineof Lespedeza juncea
var. subsessilis. The
dark green color plants
grows along to thedashed line.

autumn. In the disturbing area as river, the long time
emergence is very advantageous strategy. Then it is necessary
to complete biotope construction in summer if we hope sothat native plants may become dominant species. After
summer, the seedlings of both alien and native species
emerged for long time, from 2 to 4 months. The native plants
have an opportunity to dominate in the biotopes. The current
biotope constructions were often completed in March for theaccounting system of Japan. Then most of the new biotopes
were occupied by alien plants species. The study shows thereason of the alien plants invasion. Alien plants can easily
invade a new bared area because of the possession of theability extended germinating periods. And results alsosuggest that the construction of biotope by using“Multi-natural river-making” should be finished until
summer. When the construction was finished until summer,
there would be a same chance to invade in both alien plants
and native plants.

[8]Myers JH, Bazely DR, “Ecology and control of introduced plants,” The
Press Syndicate of the University of Cambridge, UK, 2003.
[9] Ministry of Environment, “List of regulated living organisms under theInvasive alien species act,” materials of Ministry of Environment,
2011.
[10] MacDougall AS, Boucher J, Turkington R, Bradfield GE, “Patterns ofplant invasion along an environmental stress gradient,” Journal of
Vegetation Science, vol. 17, 2006, pp. 47-56.
[11] Maskell LC, Bullock JM, Smart SM, Thompson K, Hulme PE, “Thedistribution and habitat associations of non-native plantspecies in urban
riparian habitat,” Journal of Vegetation Science, vol. 17, 2006, pp.
499-508.
[12] Masuda M, Washitani I. “A comparative ecology of the seasonalschedules for ‘reproduction by seeds’ in a moist tall grassland
community,” Functional Ecology, vol. 4, 1990, pp. 169-182.

5. CONCLUSION
1) In the river system, seeds were mainly carried when theriver rises. The seeds were carried from the embankment
slope to the high water bed part and low water bed part by therainfall.
2) In spring the amount of seedlings of alien plants was larger
than that of native plants at bared area. In autumn the amountsof seedling of both native plants and alien plants at open area.
3) There are some differences of invasion methods between
alien and native species. The strategy of alien plants islongtime seedlings emergence in spring. The strategy ofnative plants is the short time seedling emergence to survive
in spring, and the long time seedling emergence to find safe
site in autumn.
6. ACKNOWLEDGMENT
We wish to thank the member of Ministry of Land,
Infrastructure and Transport Shonai River Office for researchsupport and permission to work in the river. We also thank themembers of our laboratory for their assistance in the fieldworks.
7. REFERENCES
[1]
[2]

[3]

[4]

[5]
[6]

[7]

Shimatani Y, “Conservation and restoration of the River environment,”
Kashima-shuppan, 2000.
Van der Valk AG, Pederson RL, “Seed Banks and the Management andRestoration of Natural Vegetation,” Ecology of Soil Seed Banks, Leck,
Parker, Simpson, Eds. Academic Press, 1989, pp. 329-346.
Imahashi M, Washitani I, “Examination of possibility of a river flood
meadow restoration that uses seed bank,” Conservation Ecology
Research, vol. 1, 1996, pp.145-146.
Masuda M, “Invasion of alien plant: Status of Coreopsis lanceolata L.
in Japan,” Annual Journal of Engineering, Ehime University, vol. 21,
Mar. 2000, pp.245-251.
Washitai I, “Management of alien plants,” Conservation Ecology
Research, vol. 5, 2000, pp. 184-185.
Pheloug PC, Williams PA, Halloy SR, “A weed risk assessment model
for use as a biosecurity tool evaluating plant introduction,” Journal ofEnvironmental Management, vol. 57, 1999, pp. 239-251.
Nishida T, “Weed risk assessment-mainly in Australia and New
Zealand,” Ecology of Agriculture and Weed, Plant Species Biology
Society, Ed. Bunichisogo, 2007, pp.121-136.

442

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Recent Meandering pattern of the Irrawaddy, MYANMAR
Mayumi MATSUMOTO and Shigeko HARUYAMA
Department of Environmental Science & Technology, Mie University, JapanABSTRACT: In the lower and middle river basin of the Irrawaddy, the river is very easy to reform channels andchannel beds on the part of meandering direction during severe monsoon seasons. We measured meandering index regarding to topography, and revealed change rate of meandering index during last 60 years. We classified four zones
considering to geomorphologic units, as following; Zone A in
the middle basin of the Irrawaddy, Zone B and Zone C in thelower basin of the Irrawaddy, Zone D in delta. In our research
result, Zone A and D are stable channels, while Zone B and C
are unstable channels and easily to move the meandering
pattern.

Fig.1 Location of Myanmar
in southeast Asia
Fig.2 Location of the Irrawady
river basin in Mynanmar

Keywords: Meandering index, channel change, river gradient, Irrawaddy
1. INTRODUCTION
Flood plain formations result from the long-term cumulative
action of the flow, erosion and depositional process and themeander process will be considered in this paper. It should be
convenient that the surface geomorphology and geomorphologic process are clarified in the engineering work for theconservation the flood plain because the flow properties andsediment transport are constructed the flood plain topography.
However, in many high energy, fluvial regimes from individual floods should be so great in this paper. The influenceof recent and contemporaneous flooding in sculpting theflood plain cannot be ignored Schumm and Lichty(1963).
Even in the fluvial system, cumulative changes in channel
pattern and flood plain morphology are often great enough torequire their incorporation into engineering design andlong-term environmental changes. The evolution of channel
plan form commonly depends sensitively on local boundary
conditions. This is particularly true of flood chute development, meander cutoffs, creation and abandonment of braided-channel anastomoses. The fluvial behavior suggests thateven the retail process should be predictive for future long
period. Flood plain morphology is generally classified by the
associated channel plan form each type Lewin(1987). Nanson
and Croke(1992) also classify flood plains by the combination of sedimentary environment and formative process. The
Irrawaddy delta has the common flood plain morphology, inmonsoon Asia. Yonechi and Win Maung(1986) defined present an outline of the anastmosing river of the upper Irrawaddy and quantitative features of channel patterns. However,in the middle and lower basin of the Irrawaddy, has not been

Length2010km2

Basin415700km (Bender,1983)
13000m2/s(at Saiktha)

Discharge

max 32,600m3/s
min 2300m3/s(River regime cofficient: 14.2)
SRTM 30

SRTM 3GAIHOUZU(in 1940s)

Google earth Satellite
Imagery (in 2000)

Land relief
mapRiver gradientmap

Meandering
index

Meandering
index

Classification ofgeomorphologic units

River Transition during last 60years

clarified by channel pattern changes. The channel patterns of
this study area is classified,

443

Table.1 Profile of the Irrawaddy

has been impacted by the several channel pattern changes andheavy rainfall epoch.(Fig.1, 2 and Table.1)
(1)

(5)

2. OBJECTIVE AND METHODOLOGY

(2)
(6)

A fluvial process of the Irrawaddy river should be discussedwith the recent environmental changes caused by vegetation
changes. Therefore, we try to analyze a lateral migration in
the lower and middle basin of the Irrawaddy, where is flowing
from Mandalay intermountain basin to the mouth of river.
Therefore, we propose the following methodology, as shownin Fig.3. In our experiments, the river gradient classification
map of the Irrawaddy and land relief structure map, wereprepared by using SRTM elevation data.

(3)

(7)

(4)

(8)
(9)
(10)
(11)

えし

MyinmuMingyanPagan
Salin
Yenangyaung

Elevation(m)MagweHinthada
Nyaungdon
Knick point

Channel length(km)

(11)

(10)

(9)

(8)

(7)

(5)

(6)

Flow
direction
Lower
basin

Delta

(4)

(3)

NO.

M inor lanformFig.4 Geomorphologic units by river gradient andknickpoints(1)withValley
villageplain
namemiddlebasin
Ⅰlower basin

444

delta
river mouth

Channellength(km)Middlebasin

Gradient

(2)
(3)
(4)

Valley plainⅡ
Hilly areaⅠ
Hilly areaⅡ

898～952814～898707～814605～707

1/14,000
1/9,000
1/11,000

(5)
(6)
(7)

Hilly areaⅢ
Gorge
Small valley plain

573～605438～573366～438

1/10,000
1/12,000
1/12,000(8) Top of alluvial fan(9) Inland of delta(10) Front of delta

337～366
197～337
78～197

1/15,000
1/23,000
1/38,000(11) Estuary

0～78

1/16,000

0

(2)

(1)

Table.2 Legend(Fig.4) and river gradient

Next the GAIHOUZU provided by the Japanese army in the
1940s were prepared. Then, satellite imagery taken in 2000s
were prepared for geomorphological land classification along
the Irrawaddy. Moreover, the authors measured a meanderingindex of river condition in 1940s and 2000s, and we used thedefinition in Suzuki’s method Suzuki(1998), to quantify ariver formation and recent channel changes. Finally , wecompared the index in 1940s with the index in 2000s.
2.1 SRTM
A digital elevation model(DEM) is important ancillary data to
extract geomorphological information by remote sensing.
Especially, Shuttle Reader Topography Mission(SRTM) data
are suitable for analyzing topographical units. SRTM acquired on the flight in 2000. SRTM-3 data(90m spatial resolution) and SRTM-30 (900m spatial resolution) were used togenerate a river gradient map
2.2 MEANDRING INDEX
Suzuki(1998) defined the meandering index is expressed asfollows.
a Meandering Index MI
We measured MI between Mandalay intermountain basin andthe mouth of river, with calculation intervals in the each 5kmof the main Irrawaddy, considering to the suitable for meandering index in our study. The MI was defined by Suzuki’s
method.

3. RESULTS
3.1 Geomorphologic units
Results in our experiments are as follows. Geomorphologic
units were divided in eleven sections and eight knick points
between Mandalay intermountain basin and the river mouthfor 1,000km. In the view of the geomorphology, delta has two
knick points and inclination of river bed is the gentlest, suchas river gradient of (9)inland of delta is 1/23,000, (10)front ofdelta is 1/38,000 and (11)estuary is 0m area above sea level.
There are the knick points showing remarkable convex curvesrefer to Fig.4. In the lower basin, there are five knick
points.The knick points are showing the convex curves of theriver bed inclination in hilly area between (3)hilly areaⅠand(5)hilly areaⅢ. These areas are steeper than the other area,
especially (3)hilly areaⅠis the steepest area that gradient is
1/9,000. The river gradients in the lower basin are 1/11,000 in(4)hilly areaⅡ, 1/11,000 in (5)hilly areaⅢ, their are steep.
The inclination of other area, such as (6)gorge, (7)small valley plain and (8)top of the alluvial fan are gentler than hilly

area. The terminal knick points has concaved curve in the
middle basin on (2)valley plain Ⅱ .The river gradient of(1)valley plain Ⅰ is 1/16,000, and (2)valley plain Ⅱ is
1/14,000. The river inclination is gentle again.
3.2 Comparing the meandering index between the 1940s
and 2000s
We measured MI of 1940s and 2000s in our study. We revealed the above two MI value and eleven sections by lankingof the geomorphologic units, refer to Fig.6.We calculated theaverage of meandering index per eleven sections, and calculated change rate of MIs between 1940s and 2000s. The riverdeposition, erosion and equilibrium systems should be defined by river bed formation. These systems are importantfactor for considering to channel changes. There are four
segments between Zone A and Zone D. Zone A is (1)valley
plainⅠand (2)valley plainⅡ. Zone B is (3)hilly areaⅠ,(4)hilly areaⅡ and (5)hilly areaⅢ. Zone C is (6)gorge,(7)small valley plain and (8)top of the alluvial fan. Zone D is(9)inland of delta, (10)front of delta and (11)estuary, considering to geomorphologic units.
Zone A; Zone A is showing a erosional system. MI of(1)valley plainⅠ is 1.20, and (2)valley plainⅡ is 1.06. Thechange rate of MI between the valley plainⅠand Ⅱ, are 0.0%
and Zone A is the gentlest river slope.
Zone B; Zone B where the Irrawaddy joints the Chindwin
river that is tributary of the Irrawaddy, is high change rates ofthe MI, such as (1)hilly areaⅠ is -6.1 % and (2)hilly areaⅡis 8.0%, that are the highest degree in the other zones. Zone B
is showing a erosion and deposition system, in the same area.
Zone C; In the Zone C, the value of MI is the lowest all of thezones. The MI of (6)gorge is 1.06, (7)small valley plain is
1.08, and (8)top of alluvial fan is 0.99, because the river
channel is linear or straight. Zone C is showing a equilibrium
system. The change rates of the MI of (6)gorge is -1.09%,(7)small valley plain is 1.08% and (8)top of the alluvial fan is
0.99%, these degrees are the lowest in other zones.
Zone D; MI of (9)inland of delta is high degree of 1.29 in
1940s and 1.33 in 2000s, and change rate of meanderingindex is also high 3.0%. (10)Front of delta is most meandering area that meandering degree of 1.43. Change rate of MI of(10)front of delta and (11)estuary, are 0.7% and 0.0% in
respectively. Zone D is showing a deposition system.
4. CONCLUSION
The change rate of MIs in Zone B and Zone C are high, andthe inclinations of river bed are steep. Especially Zone B is
the highest degree of change rate and river inclination. Zone
B and Zone C have indicated that the river channel are unstable that the channel moves easily the meandering patterns,
especially Zone B is the most easily moving the meandering
pattern, because river inclination are steep. While change rateof MI in Zone A and D, are very low, and the river inclinationin gentle. In Zone A and D, change rate of MI are low, andsuggested that their channels are stable. In this study, change
rate of MI is usually affected by inclination of river bed.
Change rate of MI originated of flood occurrence.

445

5. REFERENCES[1] Harbans. Lal. Chibber(1934): THE GEOLOGY OF
BURMA, Macminllan and co, pp1-308[2] Henry. Benedict. Medlicott, William. Thomas. Blanford,
Valentine. Ball, F.R. Mallet (1965)：A manual of thegeology of INDIA and Burma, the government of India,
pp1-30[5] J. Lewin(1978): Meander development and flood plain
sedimentation: A case of study from mid-Wales, Geological Journal, vol.13 , 25-36[6] S. A. Schumm and R. W. Lichty(1963): Channel widening and flood plain construction alon Cimarron River
in southwestern Kansas, US Geological Survey professional Paper, 352-D, pp71-88[7] G. C. Nanson and J. C. Croke(1992): A genetic classification of flood plains, Geomorphology,
4, pp459-468

3.5

屈曲度 index
Meandering

3
2.5
2
1.5
1

2000年代

0.5

2000s

1940s
1940年代00
河口 of river
Mouth

200

400

600

800

1000

流路中心線(km)Channellength(km)
(11)

(10)

(9)

(7)

(8)

(6)

(2)

(5)

(1)

(4) (3)

FIg.6 Measurement results of the Irrawaddy mainstreams by meandering index
Table.3 River transition during last 60 years
Flow
direction
M iddle basin

Zone

Zone A

Lower basin
Zone B

Zone C
Delta
Zone D
River mouth

M inor lanformM eandering index(1940s)M eandering index(2000s)Change rate of
M enaderingindex(%)

Valley plainⅠ
Valley plainⅡ
Hilly areaⅠ
Hilly areaⅡ
Hilly areaⅢ
Gorge
Small valley plain
Top of alluvial fan
Inland fo delta
Front of delta
Estuary

1.20
1.06
1.15
1.12
1.25
1.06
1.08
0.99
1.29
1.42
1.13

1.20
1.06
1.08
1.21
1.29
1.04
1.11
0.97
1.33
1.43
1.13

0.0
0.0
-6.1
8.0
3.2
-1.9
2.8
-2.0
3.1
0.7
0.0

[3] Frits van der Leeden(1975)：Water Rseources of theWorld,Water Information Center,189Kimura S, J. of
Computer Science, vol. 1, Aug. 1987, pp. 23-49.
[4] A. D. howerd(1996): Modelling Channel Evolution andFlood plain Morphology, Flood plain processes, Malcolm. G.Anderson, Des. E. Walling, Paul D. Bates,
pp15-62

SystemErosion
Erosion and
Deposition
Equilibrium

Deposition[8] Fumio Yonechi, Win Maung(1986):Subdivision on theanastomosing river channel with a proposal of the Irrawaddy type, The science reports of the Tohoku University.
7th series, Geography, 36(2), pp102-113[9] Shigeru Kobayashi (2006)：The making of modern Japanin relation to East Asia countries: A perspective on the
study of Japanese military and colonial maps, E-journal

446

GEO,vol.1(1)52-66[10] Seiichiro Asano, Masashi Toyota, Satoshi Kitamura,
Goro Tomidokoro, Naoki Sugihara(2008)：The changesof river channel characteristics in the Chikuma basin bymeans of cross section drawings and aerial photographs,
Journal of Japan Applied Survy Technology, vol.15,
pp109-115[11] Ryusuke Suzuki(1998)：Kensetsu gijyutusya no tameno
tikeizu dokuzu nyuumon, kokon syoin,(2), pp223-294

447

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Physical and Chemical Properties of Tsunami Deposit in Northeast Area inFukushima Prefecture after Tohoku-Kanto Earthquake
FUJIKAWA, T., OKAZAWA, H., NAKAMURA, T., TAKEUCHI, Y. and KOMAMURA, M.
Faculty of Regional Environment Science Tokyo University of Agriculture

ABSTRACT: The damage by tsunami deposit generated by anearthquake on 11 March 2011 widely extended in the coastal area of
Tohoku distinct. From the field research and the measurements in
Minamisoma city it in northeast area of Fukushima prefecture, twokinds of tsunami deposits, sandy and muddy type deposits wereobserved. The muddy deposit whose particle size was small was
distributed the area which is relatively far from a shoreline. The
particle density of this deposit was small and the organic matter
content was very large. The sandy deposit whose particle size waslarge observed in the area near the shoreline. The particle density ofthe deposit was very large and the hydraulic conductivity of thedeposit was large although their bulk density was large. From theapproximation of the shape of the tsunami inundated area and thebulk density and the thickness of the deposits, the amount of thedeposit was estimated 124,000 m3 in volume and 193,000 Mg in dry
mass. The comparison between the two areas whose distance is
3,000 m suggested that the condition of the deposit (e.g. thethickness of the deposit) was different even if the areas wereadjoining.

Keywords: tsunami deposit, particle size distribution, particle
density, organic matter content, electric conductivity, total
amount of deposit
1. INTRODUCTION
The effect of the large tsunami, which was generated by anearthquake on 11 March 2011, extended in wide region of
Tohoku distinct, Japan. From the GIA report ([1]), theestimated area of tsunami inundated field reached
560,000,000 m2, and many farming areas and farmlands were
included in the inundated area. MAFF reported the result thatthe damaged farmlands were 240,000,000 m2 in area. Quickreconstruction and restoration of the damaged farmland havebeen strongly expected.
The effect of tsunami on farmland is classified in three
categories, (i) damage by soil salinization caused by

N 40°

N 38°

Iwate Pre.

Kashima

K-3
K-2 K-1

Mano River

MiyagiPre.
Minamisoma
Fukushima Pre.

Minamisoma city

S-2 S-1 Nitta River
Shibusathe Tsunami inundated area of 11 March 2011

Fig. 1 Research Sites in this study02,500minfiltration of sea water into the soil, (ii) deterioration of thesoil physical, chemical and biological properties bydeposition and (iii) erosion of surface soil by tsunami ([2]),and these effects are combined in some cases. The damage bytsunami deposition would be large especially in the coastal
area. To select the method of reconstruction of the damaged
farmland by tsunami deposit, information about not only theamount of the deposits but also its properties is considered tobe very important. The change of the properties of the soil
below the deposit is also needed to be understood before thetreatment of damaged farmlands because the deposition ofearth and sand could compress the surface soil of the
farmland and the chemical materials eluted from the deposit
could infiltrate into the soil,
The purpose of this study is to clarify physical and chemical
properties of tsunami deposit and farmland soil below it innortheast area in Fukushima prefecture where the damage ofthe tsunami brought by the Tohoku-Kanto earthquake wasserious.
2. SITES AND METHODS
2.1 Research Sites
The field research was carried out in two areas, Kashima andShibusa in Minamisoma city located in northeast part of
Fukushima prefecture (Fig.1). The shoreline of these areas
runs north and south, the wave direction of tsunami was west.
The tsunami reached 3,200 m in Kashima and 2,500 m in
Shibusa. Three sites in Kashima (K-1, K-2 and K-3) and twosites in Shibusa (S-1 and S-2) were selected for themeasurements. These sites had been used as paddy fields
before the tsunami event. At each site a sampling pit, which is0.35 m (K-1 and S-1) or 0.25 m deep, 0.50 m wide and 0.50 m
long, was dug to investigate the profile of the deposit and thesoil.
The field survey was conducted on 17 June 2011, 3 months
after the tsunami event. The total amount of precipitation
between the tsunami event and the survey was calculated tobe 230 mm from AMeDAS data in [3].
2.2 Sampling and Measurements
Thickness of the deposit was measured by observation of the
profile of the sampling pit. Disturbed and undisturbed
samples were taken from the profile. The undisturbed sample
was taken by the 100 cm3 core sampler with 5.0 cm indiameter and 5.1 cm in height.
Particle size distribution was determined by hydrometer
method (< 0.106 mm) and sieving (> 0.106mm) after
decomposition of organic materials by hydrogen peroxide447

(referring JIS A 1204). Water content and bulk density of thesample was measured by oven dry (105 °C, 24 hours) with the
disturbed and undisturbed sample, respectively. Particle
density of the disturbed sample was determined bypycnometric method (JIS A 1202). Ignition loss, whichindicated the content of organic matter, was calculated fromthe difference between the weight of the oven dried sample
and that of the sample heated to 800 °C for 3 hours in a muffle
furnace. Saturated hydraulic conductivity of the undisturbed
sample was determined by falling-head method. A water
extract was obtained from 4 g of the sample with 20 gdeionized water (1:5 method) and electric conductivity (EC)of this extracts was measured. Concentrations of chloride ion
and sulfuric ion were analyzed by IC (detector: CDD-10Avp,
SHIMADZU) for the extracts filtrated by 0.22 m filter. The
measurements of particle size distribution, EC and anion
concentration were made for the samples only in Kashima.
The replication of the measurement was one for particle size
distribution, EC and anion concentration and two for otherproperties.
3 RESULTS AND DISCUSSION
3.1 Thickness of the Tsunami Deposit
In Kashima the thickness of the tsunami deposit decreased
with distance from the shoreline as some literature ([4]-[6])reported (Table 1). Muddy deposit was observed in at K-3
whose distance from shoreline was most large. The
observations of such muddy deposit (such as, surface muddy
deposit layer or muddy caps in [7] - [9]) have been also
reported. The thickness of the deposit in coastal area became
larger because the large fall of velocity of the tsunami in thisarea promoted the deposition of including earth and sandcarried from sea bed. In Shibusa, on the other hand, thesurface deposit layer of S-2, which is closer to the shoreline,
was thicker than that of S-1. Some reports ([4],[10]) observed
no clear relationship between the distance from a shoreline
and the thickness of tsunami deposit and it is explained interm of the change in water flow velocity and direction by
microtopography ([5], [11]), deposition and erosion by theoutflow or the second and third inflow ([12], [13]). When thetsunami ran up to the field so fast and the water volume oftsunami is so large, the deposition may not occur at thecoastal area because of its large transport ability. The
thickness at S-1 was different from the average of those of
K-1 and K-2 although the distance of S-1 from the shoreline isclose to the average of those of K-1 and K-2. It is suggested
that the condition of the deposit considerably largely varies
between two areas whose distance is only 3,000 m.
3.2 Particle Size Distributions of the Deposit in Kashima
The particle size distributions of the surface deposit (0 – 0.05
m depth) at each measurement site in Kashima shown in
Fig.2(a). The deposit at K-3, whose distance from theshoreline is largest, was smaller (D50 = 0.04 mm) than those at
K-1 (D50 = 0.2mm) and K-2 (D50 = 0.3 mm). And especially,
the sand fraction at K-3 was remarkably smaller. However,the particle size of the deposit of K-2 was larger than that of
K-1, and a clear relationship was not observed between the
distance from the shoreline and the particle size. These resultsof the relationships between the distance and the particle size
agree with that obtained in [9], [14]. In [14] it was found thatthe coarse fraction (very coarse and coarse sand) of thedeposit decreased with the distance although the range of thefine fraction (fine sand) with the distance was narrow. It isconsidered to be effective to clarify the spatial variability ofthe particle size classified into two or three parts (e.g. course
fraction, fine fraction and muddy fraction).
The particle size of the deposit increased with depth at K-1 as[8] and [9] reported. This particle size distribution in the
direction of depth is considered to be a result of the differenceof the settling velocity related with the particle size as shownin Stoke’s law. This particle size distribution in the directionof depth would become clearer with the passage of the time
because the surface deposit will roll up and settle again andagain in the ponding condition after heavy rain. The deposit at
K-1 surface included about 10 % of fine fraction (< 0.1 mm),which suggested that the muddy deposit was mixed in thedeposit at coastal area even if the deposit was judged to besand deposit by observation.

Table 1 Location and thickness of the tsunami deposits at the research sites
Kashima

Shibusa

Site name

K-1

K-2

K-3

S-1

S-2

Longtitude

37゜41 ´29″37゜41 ´49″

37゜41 ´59″

37゜38 ´15″37゜38 ´28″

Latitude141゜ 0 ´28″

140゜59 ´46″

140゜58 ´51″

141゜ 0 ´56″

140゜59 ´55″

Distance fromshoreline (m)

300

1300

2700

700

2200

Thickness of deposit (m)maddy deposit
sandy deposit
0.25

0.10

0.05
-

0.05

0.05
0.05

448

Table 2 Water content, densities and ignition loss of the deposits

100

Bulk
Sampling Deposit/ Water
Particle Ignition
depth* soil type** content density densityloss(%) (Mg m-3) (Mg m-3)(m)(%)(a) 0-5 cm
80

Accumulative mass fraction of particles smaller than d (%)

60
40
Site
K-1
K-2
K-3

20
0
10 -2
100

10 -1

10 0

10 1

Kashima0-0.05
K-1 0.10-0.15
0.20-0.25
0.30-0.35

sandy
sandy
sandy
paddy

10.0
13.0
13.9
26.3

1.59
1.69
1.66
1.24

2.95
2.93
3.03
2.61

2.1
0.7
0.9
9.8

K-2

0-0.050.10-0.15
0.20-0.25

sandy
paddy
paddy

3.6
27.3
24.6

1.35
1.21
1.36

3.01
2.55
2.61

1.5
11.0
8.5

K-3

0-0.050.05-0.10
0.20-0.25

muddy
paddy
paddy

24.6
25.8
27.5

0.87
0.97
1.12

2.24
2.51
2.59

25.6
12.1
12.2

0-0.050.10-0.15
0.20-0.25
0.30-0.35

sandy
paddy
paddy
paddy

7.7
31.5
26.4
13.7

1.28
1.15
1.28
0.99

2.79
2.49
2.45
2.70

3.6
13.0
13.9
17.0

0-0.050.05-0.10
0.20-0.25

muddy
sandy
paddy

13.7
23.4
17.0

0.90
1.14
1.03

2.56
2.65
2.59

19.6
6.7
11.1

Sibusa(b) K-1
80

S-1

60
40
20
0
10 -2

S-2

Depth (cm)0-2.5
10-12.5
20-22.5
30-32.5

10 0
10 -1
Diameter of particles d (mm)* depth from the surface of the deposit** sandy : sandy deposit, muddy : muddy deposit,
paddy : paddy soil before tsunami inundation

10 1

Fig. 2 Particle size distributions of deposit

3.3 Water Content and Bulk and Particle Densities of
the Deposit
The water content of the deposit tended to be smaller thanthat of the paddy soil below the deposit (Table 2). Especiallythe sandy deposit in the surface layer was dry, of which watercontent was about 10%. The bulk density of the sandy depositof K-1 surface layer was 1.6 Mg m-3, which is 1.8 times aslarge as those of the muddy deposits at K-3 and S-2 (about 0.9
Mg m-3). The increase of the bulk density was not confirmed
from the measured values of the bulk density and the effect ofcompaction by the deposit on the farmland surface soil was
not recognized since. It is considered that the influence ofcompaction by the deposit is comparatively small because thesoil type of the surface of paddy fields in this area is heavy
clay and the drainage before the tsunami event had alreadyconsolidated the surface soil. On the other hand, the influenceof compression by the deposit can increase with time becauseof the iteration of drying and wetting ([15]). Since bulk
density of soil, which is related to the water infiltration andretention, is one of the important properties for the agriculture,

further measurements are needed to clear effect of thecompaction of surface soil by the tsunami deposit.
The particle density of the particle of a sandy deposit was 2.8- 3.0 Mg m-3, which was larger than that of general soil offarm land in Japan (2.5 - 2.7 Mg m-3), while that of muddy
deposit was smaller (2.2 Mg m-3). The large particle densityof the sandy deposit can be due to the content of metal or
heavy metal. The content of cadmium and chrome in thedeposit sampled after the 2004 Indian Ocean tsunami isreported in [16], [17]. They insisted that the source of theheavy metal was not only lithogenic but also anthropogenic in[17], in other word, the metal content of the deposit varies
with the condition of industry of the neighbor areas. The
analysis of the content of metal of the deposit is necessary
also in this area. The small particle density of muddy deposit
is interpreted as the high content of organic material. The
ignition loss of the muddy deposit was 25 % and higher thanthose of sandy deposit and paddy soil (Table 2). The organic
matter included in the muddy deposit could contain the
microorganisms living in the ocean. Some kind of the
microorganisms may have toxicity against human health.
Microbiological analysis of the deposit will be expected
before the removal and disposal of the deposit to restore thefarmland. The ignition loss of the surface deposit at K-1(0-0.05 m) was higher than that of the deposits collected fromdeeper layers (0.10 - 0.15, 0.20 - 0.25 m) at K-1. This result449

supports the discussion that the deposit in surface layer of K-1
included the muddy component even if that site is closer tothe shoreline.
3.4 Hydraulic conductivity of the deposit
The saturated hydraulic conductivity of a sand deposit was
10-4 - 10-3 m s-1, which was about ten times as large as that of
surface soil of paddy field (Fig.3). The relative large value ofthe saturated conductivity of the paddy soil may be due to themixing of organism materials (compost, rice straw, etc.) or
sandy soil brought from another place for the improvement ofphysical properties of the surface soil. The hydraulicconductivity of the muddy deposit was, conversely, smaller
than that of the sandy deposit, and that of the deposit at theS-2 surface was about 10-7 m s-1. Since the bulk density of the
muddy deposit was very small, the pore volume of the deposit
was thought to be rationally larger even if the particle densityof it was small. This contradiction can be explained by thesmaller pore size of the muddy deposit which is formed bysmall particles included in the deposit. The hydraulicconductivity of K-3 was smaller than that of sandy deposit but
10 times large as that of the deposit of S-2 surface. There weresome cracks brought by drying process on the surface at K-3
and there is a possibility that the downward infiltration of
rainfall water is promoted through such surface cracks.
3.5 EC and Cation Contents of the Deposit in Kashima
The value of electric conductivity of the deposit sampled in
Kashima was 0.39 - 0.13 S m-1 for the sandy deposits andabout 1.3 S m-1 for the muddy deposit (Table 3). The EC
value of muddy deposit at K-3 was very large. These valuesof EC are relatively higher compared to that of the paddy soil
under the deposit (0.04 - 0.29 S m-1). The EC of the paddy soil
below the deposit may be increased by the infiltrated water-0.3
Depth from the soil surface
before tsunami invasion (m)

sandy

-0.2
Depositwhich contented the high salinity affected by the deposit.
From the result of the analysis by IC, both Cl- and SO42concentrations of the extracted water decreased with depth.
The ion contents with dry material were estimated by thewater content of the deposits and the ion concentrations of theextracted water. The estimated Cl- and SO42- contents ofsandy and muddy deposit were 0.5 % and 0.05 % for sandy
deposit, 3 % and 0.2 % for muddy deposit, respectively.
These estimated value agree with the result of [17], which
reported the Cl- and SO42- content of the tsunami deposit are
0.03 - 3.4 % and 0.05 - 3.5 %, respectively.
In [2] it is described that salt leeching is needed when the Clcontent is larger than 0.1 % for paddy field and 0.05 - 0.07 %for upland field. In this area the contents of both the deposit
and the soil are exceeded that standard, and leeching process
will be needed before the field is used as a agricultural land
when the removal of the deposit has completed.
3.6 Estimation of the Amount of the Deposit in Kashima
The total amount of the deposit in Kashima was estimated by
the measured value of the depth and the bulk density of thedeposits (Fig. 4). The shape of the area, the bulk density of thedeposit and the relationships between the distance from theshoreline and the thickness of the deposits were approximatedby the assumption described below.
1) This area is approximated by the trapezoid whose upper
base is 2,100 m, bottom base is 4,500 m and height is 2,900m.
2) The bulk density of the deposit is assumed 1.66 Mg m-3 forsandy deposit and 0.88 Mg m-3 for muddy deposit,
respectively.
3) The thickness of the deposit is expressed by the function of
only the distance from the shoreline. And the thickness ofthe sandy deposit decreased and that of the muddy deposit
increased linearly with the distance.
4) Only the sandy deposit exited at K-1 (300 m from theshoreline) and the muddy deposit at K-3 (2,700 m) and thethickness of these deposits are 0.25 m at K-1, 0.05 m at K-3,
Table 3 EC and Cl- and SO4 2- contents of the deposits

muddy

-0.1
Sampling
depth (m)00.2

0.3 -8
10

10 -6

10 -5

10 -4

10 -3

K-2

K-3

S-1

S-2

Fig. 3 Distributions of saturated hydra ulic conductivity

450

SO42Cl-1(mg g ) (mg g-1)

1070
750
750
460

422
82
80
33

5.6
4.5
4.2
3.0

2.2
0.5
0.4
0.2

K-2

0-0.050.10-0.15
0.20-0.25

0.21
0.13
0.04

590
410
100

57
50
4

3.9
2.2
0.6

0.4
0.3
0.0

K-3

0-0.050.05-0.10
0.20-0.25

1.28
0.29
0.14

4830
680
370

309
132
65

28.4
4.5
2.3

1.8
0.9
0.4

Saturated hydraulic conductivity(m s-1 )
K-1

ClEC
SO 42-1(S m ) (ppm) (ppm)
0.25
0.39
0.25
0.17

K-1

10 -7

Dry soil base0-0.050.10-0.15
0.20-0.25
0.30-0.35

0.1
Paddy
soil

Extracted waterrespectively.
5) The deposit at K-2 surface (0 - 0.05m) was assumed to bea mixture of the sandy and muddy deposit. The mixing
ratio was calculated by interior division of the assumed
bulk density of sandy and muddy one and the measured
bulk density of the deposit at K-2 surface (0 - 0.05m). The
mixing ratio of the sandy deposit at K-2 was calculated tobe 0.60 and the thickness of the sandy deposit in the surface
was determined 0.03 m and that of muddy deposit was 0.02
m. The deposit of 0.05 – 0.10 m at K-2 is assumed to be thesandy deposit.
The relationship between and the distance from the shoreline(x m) and the width of the area (Y m) was obtained:Y  0.828 x  4500

(1)from the assumption 1) and relationships between x and theamount of the thickness of the deposit (y1 m: sandy, y2 m:
muddy) :(a) The approximation of the shape of Ka shima
2,900

4,500

2,100

K-3

K-2
K-1

actual inundated
area

idealized area for
estimation(b) The a pproximation of
thickness of the deposit
muddy deposit(0.88 Mg m-3)

0.10

0.25

K-2
(1300)(x < 1770)(2)y 2  0.0000021x  0.08 (x < 370)

(3)from the assumption 3) – 5).
By multiplication of the area and the thickness of the deposit,
the total volume of the deposit was estimated 124,000m3(sandy deposit: 106,000 m3, muddy deposit: 18,000 m3). The
total dry mass which was estimated by the multiplication of
the volume and the bulk density of the deposit was
193,000Mg (sandy deposit: 176,000Mg, muddy deposit:
17,000Mg). To be the estimation result more precisely,
further measurements of the thickness and density of thedeposit in the area. The information of wave height and speedof tsunami which affect the amount of tsunami deposit iseffective to estimate the amount of the deposit moreaccurately ([13]). On the other hand, no clear tendency
between the distance and the thickness was observed in
Shibusa. To estimate the total amount of the deposit in thisarea it is necessary to investigate more about the
microtopography of this area and the sea base in front of thisarea because not only microgeographical features of theground but also the spatial variability of properties of seabedcan affect the amount and the properties of the deposit.
After the removal of the deposit, salt leeching is needed
before the planting of the damaged farmland ([2]). If theamount of the deposit is so small that the change of thechemical and properties is negatively small, the mixture of
the deposit into the surface soil and upside down tillage areavailable to restore the farmland. The Cl- content of the paddy
soil at K-2 (0.20 – 0.25m) was smaller than the standard of theleeching. At this site, only the removal of the deposit can beenough for the restoration of the farmland without leeching.
But it is also reported that the peak of the ion concentration
moves downward ([10], [17]) with time. The removal of thedeposit including toxic substances is expected to becompleted as soon as possible to prevent the contamination of
the soil below the deposit.
4 CONCLUSION

sandy deposit(1.66 Mg m-3)

K-3(2700)y1  0.00017 x  0.30

K-1(300*)(m)*: the distance from the shoreline

Fig. 4 Schema tic diagram for estimation of
the amount of the deposit

To clarify the physical and chemical properties of the tsunami
deposit caused by Tohoku-Kanto earthquake on 11 March
2011, the field research and the measurements of theseproperties were done in Minamisoma city located in northeast
region of Fukushima prefecture. Two kinds of deposits, sandy
and muddy type deposits were observed. The sandy deposit
whose particle size was large observed in the area near theseashore. The maximum thickness of the deposit was 0.25 m
and decreased with the distance from the shoreline. The
particle density of the deposit exceeded 2.8 Mg m-3, whichindicated the possibility of the content of metal and / or heavy
metal in it. The hydraulic conductivity of the deposit waslarge although their bulk density was large. The rapid
infiltration is likely to promote the downward transfer of451

these toxic materials into the farmland soil. The muddy
deposit whose particle size was small and the density was
small is distributed the far area from a shoreline. The
maximum thickness of the deposit was 0.05 m. The particle
density of this deposit was small and the organic matter
content was very large, which suggested that the dried andsmall particle of the deposit injured human withmicroorganism or its products including in the deposit. The
cation concentrations of the extract water of this deposit
absorb was large which indicated the possibility that the anion
absorption of it was also large. From the approximation of the
shape of the area and the bulk density and the thickness of thedeposits, the amount of the deposit was estimated 124,000m3
in volume and 193,000 Mg in dry mass. The thickness of thedeposit in Shibusa whose distance from Kashima is about
3,000 m, however, did not increase with the distance from theshoreline and it is suggested that the condition of the deposit
was different among the location.

[17] Szczuciński W., et al., “Effects of rainy season on mobilization ofcontaminants from tsunami deposits left in a coastal zone of Thailand
by the 26 December 2004 tsunami”, Environ. Geol., vol. 53, 2007,
pp.253-264

5 REFERENCES[1]

The Geospatial Information Authority of Japan (GIA), “About the totaltsunami inundated area (rough estimated value) (5th report)",http://www.gsi.go.jp/common/000059939.pdf , 2011 (in Japanese)[2] Ministry of Agriculture, Forestry and Fisheries (MAFF), Manual forsalt leaching",
http://www.maff.go.jp/j/press/nousin/sekkei/pdf/110624-01.pdf “,
2011 (in Japanese).
[3] Japan
Meteorological
Agency(JMA),“AMeDAS",
http://www.jma.go.jp/jp/amedas_h/, 2011 (in Japanese)[4] Nishimura Y. and Miyaji N., “Tsunami deposit from the 1993
southwest Hokkaido earthquake and 1640 Hokkaido Komagatake
Eruption, Northern Japan", Pegeoph, Vo. 144, 1995, pp.719-733[5] Peters R., et al., “Distribution and sedimentary characteristics oftsunami deposits along the Cascadian magin of western North
America", Sed. Geol., vol 200, 2007, pp.372-386l[6] Matsumoto D., et al., “Thickness and grain-size distribution of 2004
Indian Ocean tsunami deposits in Periya Kalapuwa Lagoon, eastern Sri
Lanka", Sed. Geol., vol 230, 2010, pp.95-104[7] Richmond B. M., et al., “Deposit, flow characteristics, and landscapechange resulting from the September 2009 South Pacific tsunami in
Samoan islands", Earth-Sci. Rev., vol. 107, 2011, pp.38-51[8] Morton R. A., et al., “Physical criteria for distinguishing sandy tsunami
and storm deposit using modern examples", Sed. Geol., vol 200, 2007,
pp.184-207[9] Gelfenbaum G. and Jaffe B., “Erosion and sedimentation from the 17
July, 1998 Papa New Guinea Tsunami", Pure appl. Geophys., vol. 160,
2003, pp.1969-1999[10] Mcleod M.K., et al., “Soil salinity in Aceh after the Decmber 2004
Indian Ocean tsunami", Agric. Water Manag., vol. 97, 2010,
pp.605-613[11] Dawson A. G. and Stewart I., “Tsunami deposit in the geologicalrecord", Sed. Geol., vol 200, 2007, pp.166-183[12] Nakayama F. and Shigeno K., “Inflow and outflow facies from 1993
tsunami in southwest Hokkaido", Sed. Geol., vol. 187, 2006,
pp.139-158[13] Goto K. and Fujino K., “Problems and perspectives of the tsunami
deposits after the 2004 Indian Ocean tsunami, Jour. Geol. Soc. Japan,
vol. 114, No.12, pp.599-617 (in Japanese with English abstract)[14] Moore A. L., et al., “Landward finding from multiple sources in a sandsheet deposited by the 1929 Grand Banks tsunami, Newfoundland",Sed. Geol., vol 200, 2007, pp.336-346[15] Hulugalle N. R., et al., “Physical properties of tsunami-affected soils in
Aceh, Indonesia: 2 1/2 years after the tsunami", Catena, vol. 77, 2009,
pp.224-231[16] Szczuciński W., et al., “Contamination of tsunami sediments in scoustal zone inundated by the 26 December 2004 tsunami in Thailand”,
Environ. Geol., vol. 49, 2005, pp.321-331

452

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Changes in Bottom Ssediment Caused by Construction of the Airport Island in
Ise Bay, Japan.
Maki OYAGI1, Motohiro KAWASE2 and Akihiko YAGI1
1

Dept. of Engineering, Aichi Institute of Technology
Dept. of Human Science, Aichi Mizuho University2


ABSTRACT: In order to investigate the impact on the marine
environment from the construction of the Chubu International
Airport Island in Ise Bay, Japan, we made careful observations toclarify the actual seawater, bottom sediment condition and thecomposition of benthic fauna at some stations located from east to
south of the airport island. Grain size compartment, moisture
percentage, ignition loss, carbon, nitrogen, phosphorus and sulfur
were measured. The ignition loss values in surface sediment in 2003
averaged 150% times higher than that of data in 2007. The benthic
fauna revealed the species and number of individuals were very few.
Organic matter increases, the species structure of the benthos ismade poor. The average number of individuals and species
compared before and after the Airport construction clearly showed aremarkable decrease after the construction. This phenomenon mightreflect the ongoing progress of eutrophication and oxygen depletionaround the airport island sea.

2. Methods
2.1 Study site
The study site was located on the Ise Bay sea area between
the Chubu International Airport Island and Chita Peninsula.
Seven sampling points were selected: very shallow sea area;
A1 (4 m depth), A2 (3 m depth), A3 (5 m depth), A4 (7 m
depth) and shallow area; B1 (11 m depth), B2 (11 m depth),B3 (13 m depth), where water quality and sea sediment were
examined. Observations were conducted twice a year (2002,
2004, 2005, 2006 and 2007) and three times in 2003,
respectively.

Keywords: Bottom sediment, Airport Island construction,
Carbon, Nitrogen, Benthic fauna

1. INTRODUCTION
The Chubu International Airport Island was constructedalong the east coast of Ise Bay where fishing grounds havebeen especially excellent. However, the Oceanographic
Soiety of Japanese Environmental Issues Commission
warned that this sea area environment would be negatively
impacted by constructing the airport island [1]. On the other
hand, the Central Japan International airport Co., Ltd. andAichi prefecture reported that an environmental assessment inconnection with the airport island construction revealed
minor damage to fishing grounds [2]. After the airport island(580 ha) was built, the other side area (130 ha) of the airport
island was developed. After the airport opening in 2005, theinvestigation near the airport island has been continually
carried out every year by Aichi Prefecture and the Central
international airport Co., Ltd. [3]-[5]. Originally, theTokoname City open sea is going sea area, while the fresh
water in the Kisosansen River is mixed with seawater, whichreading tends to stagnate on the south side in the airport
island.
When diving fishermen who have been enaged in fishing inthese sea areas, were asked why the shellfishing decreased soshaphy recently, they agreed to coorerate with the present
study by assisting with the core sampling of the bottom
sediments. This study the purpose of our study is to clarify
the effect of large structure of the coastal zone on the nearby
sea area environment for water quality, bottom sediments andthe benthos.
Fig.1 Study site453

2.2 Sampling and analysis
Sediment core samples, three acrylic pipes (5 cm in diameterand 50 cm in length), were taken from the each location withtwo divers, and two core samples of three cores were
subdivided into 1 to 2.5 cm-thick sections with a knife by
extruding sediment core vertically.
Ignition loss (I.L.: 650℃), total nitrogen (TN), total organiccarbon (TOC) and total sulfur (TS) were analyzed with a
CHNS analyzer (Perkin Elmer CHNS/O analyzer 2400). The
grain size was analyzed using the test method for particle size
distribution of soils (JIS A 1204) and was fractionated to six(75 μm or less, 75 to 125 μm, 125 to 250 μm, 250 to 500 μm,
500 to 2 mm, 2 to 4 mm).
Bottom sedimentation rate was estimated by applying the210
Pb method using the bottom sediment cores sample in 2004
and was 0.5 cm y-1 in the surrounding area [7]. Furthermore,
0.06〜0.76 g cm-2 y-1 of the sedimentation rates were reportedby the applying 210Pb method of ten stations in Ise Bay [8].
Therefore, it was necessary to analyze sediment sample in
1-cm to 2.5-cm-thick sections vertically in order to clarify theenvironmental change around the airport island.
Macrobenthos was collected by the quadrat method (25 cm
square, 25 cm depth), and the sediment was separated by asieve (φ1 mm). The kind and number of the mollusks wereobtained.

3 RESULTS AND DISCUSSION
3.1 Particle size distribution.
Organic matter contents of in B1, B2 and B3 were highratio of silt-clay fraction and those of B1, B2 and B3 were39-40%, 31-41% and 20-28%, respectively (Table 1). The
silt-clay fraction was big for B2 and B3 in 0-1 cm layer in thepole surface sediment, and with deepening form 1-2 cm layer
to 2-3 cm layer this fraction became clearly small tendency.
In spite of shallow A1, the silt-clay fraction was relatively
large with 11-13% and then it was indicated that the sea water
was easy to stagnate.

3.2 Ignition loss, carbon, nitrogen, phosphate, sulphur
In case of the mean of surface layer (0-1 cm) for carbon andnitrogen ratio (C/N) from September, 2004 to October, 2005,
the values of more than 10, less than 7 and the middle were
stations of B2 (11)- B3 (13)- A4 (16), stations of A2 (7)-A1(6) and stations of B1 (9)-,A11 (9), respectively. From thisfact, the decomposition of the organic substance seems toprogress at sites (B2, B3), where the depth is great, and thesite A4 in the pole shallow sea area. At the site with theshallow depth, the influence or the comparatively new
sediment was strong, suggesting that B1 and A2 were in
between. As for the C/N value, on the other hand, in light of
the result from September 2004 to October 2005, a clear
difference was recognized between the surface layer ( 0-1 cm)
and the layer beneath. The C/N value has become smaller,suggesting that, under autochthonous production, thesediment seems to immediately deposit on this very surface,causing at least a temporary disturbance with a typhoon. The
sites where the stratified structure was clearly recognized
were B1, B2, B3, A3 and A4, and A1 and A2 estimated thatthe disturbing ( about 0～7cm ) of sedimentary layer occurs
in the shallow site coastal from the vertical distribution of
C/N,In order to show the poor oxygen condition of the bottom
sediments, the total sulfur was determined.
The ignition losses for B1, B2 and B3 with great depth
were high concentrates and the total sulfur was also high withthe maximum value of 1～5 mg g -1 ( Fig. 2). The total sulfurof the surface in the pole surface of 0～2 cm layer tended tobe high. Furthermore, the value total sulfur of about 0.3~0.7
mg g-1 , of total sulfur was measured in the relatively shallow
sea area of A4, where bottom mud evidenced a hydrogensulfide smell.
In addition, red tide was appeared in the investigation sea
areas in July 2005. All bottom core samples to the laboratory
turned black within around 6 hours. It showed that aremarkable reduction condition had occurred in that bottom
area.

Table 1 The particle size distribution
stationdepth (cm)75μm>75-125μm

125-500μm

500μm-2mm

2-4mm

>4mm

A1

0-1
1-2
2-3
0-1
1-2
2-3
0-1
1-2
2-3
0-1
1-2
2-3
0-1
1-2
2-3
0-1
1-2
2-3
0-1
1-2
2-3

4.52
5.75
6.34
2.72
3.11
2.96
3.13
2.69
2.95
1.80
1.76
1.77
30.21
24.10
26.97
38.29
22.66
18.66
16.87
14.22
13.94

5.89
7.17
4.81
0.87
0.88
0.95
2.11
1.30
1.37
0.30
0.00
0.00
9.06
10.48
13.73
7.73
12.11
12.29
11.88
9.78
6.40

87.50
85.47
85.64
92.07
90.72
90.50
79.92
75.15
72.91
85.89
83.18
82.41
44.80
50.60
45.55
35.33
49.92
47.83
66.76
70.17
73.72

2.10
1.30
2.82
3.04
3.36
4.92
14.84
20.40
20.92
11.16
13.93
13.32
8.30
9.24
6.66
14.14
12.80
12.86
3.43
3.70
3.22

0.00
0.32
0.39
0.74
0.37
0.68
0.00
0.45
1.86
0.86
0.54
2.50
3.59
4.81
3.43
1.75
1.96
2.31
0.73
1.69
1.63

0.00
0.00
0.00
0.56
1.57
0.00
0.00
0.00
0.00
0.00
0.59
0.00
4.05
0.77
3.66
2.75
0.55
6.04
0.33
0.44
1.09

A2
A3
A4
B1
B2
B3

454

%
3.3 IL Comparison around the airport construction
Ignition loss in 2003 and 2007 in the surface layer (at 0-2.5
cm depth) from the result of the dating which corresponds toabout 5 years was compared at each point (Fig. 3). The 2003
value corresponds from 1999 to 2003, and the 2007 value tothat from 2003 to 2007. This 5 cm thickness was equivalent
to approximately 5 years from a result of the measurement of
the sedimentation rate. At deep areas of B1, B2 and B3, the2003 value is remarkably higher than in 2007, averaging
150%. 112% was obtained at A2, A3 and A4 site in very
shallow sea. Beside, TS4 data showed the most remarkably
changed value of 190% despite the shallow depth. It
appeared at a place especially where the influence on the deepbottom sediments of the construction airport island started in
November, 2000 was remarkable.
In this way, the progress of the bottom mud pollution clealychanged at A1 in very shallow sea and also at A4 where thesea bed had a convex shape.

station

0

5

0-

2-

2-

4-

4-

6-

6-

9.5-

9.55

10

0-

0-

2-

2-

4-

4-

0

5

0-

2-

2-

4-

4-

6-

6-

9.5-

9.50

5

0

5

10

5

10

5

10

0

10
02-

2-

（ｃｍ）

10

0

10

0-

0-

4-

4-

6-

6-

9.5-

9.50

depth

5

9.5-

9.5-

5

10

0-

0-

2-

2-

4-

4-

6-

6-

9.5-

9.5-

0

5 REFERENCES

0

6-

6-

The authors wish to thank the Himaka Island Fishermen’s
Union for their valuable field assistance. This study was
supported by the 13th Pro Nature Fund, a grant forJapan-Related Research Project 2006 from the Sumitomo
Foundation, the Aoi Umi fund 2009 from the Oceanographic
Society of Japan.

455

10

0-

0

4 ACKNOWLEDGMENT[1] The oceanographic society of Japanese environmental
issues commission “For construction of artificial islands
Central International Airport” (in Japanese), Journal of
Oceanography, vol.8, 1999, pp. 349-357.
[2] Central Japan International air port Co., Ltd. and Aichi
prefecture “The environment assessment’s brief about theconstruction project of the Central Japan International
Airport and the constructed project of the airport island area
land for development (abstract)” (in Japanese) , 1998,
269pp.
[3] Central Japan International airport Co., Ltd. and Aichi
prefecture “Annual report of environmental monitoring
results (2002) about the construction project of the Central
Japan International Airport, the constructed project of the
airport island area land for development and the constructed
project of the other side of the airport island ” (in Japanese),
2003, 442 pp.
[4] Central Japan International Airport Co., Ltd. and Aichi
prefecture “Annual report of environmental monitoring
results (2003) about the construction project of the Central
Japan International Airport, the constructed project of the
airport island area land for development and the constructed
project of the other side of the airport island” (in Japanese),
2004, 426 pp.
[5] Central Japan International Airport Co., Ltd. and Aichi
prefecture “Annual report of environmental monitoring
results (2004) about the construction project of the Central
Japan International Airport, the constructed project of the
airport island area land for development and the constructed

%

5

0

5

10

0

5

10

0

5

10

10

0-

0-

2-

2-

4-

4-

6-

6-

9.5-

9.5-

0

5

10

0-

0-

2-

2-

4-

4-

6-

6-

9.5-

9.5-

2004.7

2004.9Fig.2 Ignition loss in vertical distribution

%

10

5

0
A1

A2

A3

A4

B1

B2

B3

Fig.3 Ignition loss of surface layer (0-2.5 cm
depth) in 2003 (white bar) and in 2007 (black bar)

project of the other side of the airport island” ( in Japanese),
2005, 464 pp.
[6] The ministry of the environmental water quality beau,“Marine environmental monitoring guidelines”, The
ministry of the finance Printing beau, 1997, 136pp.
[7] Saijo Yatsuka, Hisayoshi Terai, Maki Umemura, Akihiko
Yagi, Mariko Nagano, Yoshihisa Kato, Motohiro Kawase,
Yasuo Matsukawa and Katsuyuki Sasaki, “Degradation ofseawater and sediment quality and benthic faunacomposition caused by construction of the Chubu
international airport island in Ise bay, Japan” (in Japanese),
2008, Oceanography in Japan, 17(4), 281-295.
[8] Lu, X. and E. Matsumoto (2005): Recent sedimentation
rates derived from 210Pb and 137Cs methods in Ise Bay, Japan.
Estuarine, Coast. Shelf Sci., 65, 83-93.

456

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Proposal of simple measurement method for evaporation rateby using oxygen isotopic ratio in the Inawashiro Lake
Satoshi Miyahara1), Hiroyuki Ii2)
1)Graduate School of Systems Engineering, Wakayama University
2)Faculty of Systems Engineering, Wakayama University



ABSTRACT: The purpose of this study is to develop simple
method for estimating evaporation rates of water using oxygenisotopic ratio. From the laboratory test, strong negative correlation
between average humidity and δ18O change per unit evaporation
rate was observed and humidity was clarified to be importantparameter. We can calculate evaporation rate from the amount ofchange of the δ18O during evaporation, the amount of change ofthe δ18O per unit evaporation rate estimated empiricallydetermined under each humidity condition and humidity.
The amount of evaporation per year in the Inawashiro lake inFukushima was estimated to be 670 mm/year from the relation
between humid and evaporation rate and the estimated value wasin agreement with the calculation result of previous research,
600mm/year. Therefore, the simple estimation method is effective
for estimation of evaporation rate of an actual lake or a pond
especially in a dried area because the measurement error is small
under the low humid condition.

Keywords: evaporation rate, oxygen isotopic ratio,
humidity, the Inawashiro lake

Recently, water shortage is caused by climate change such
as Global warming. Therefore, estimation of amount of
evaporation for dam, reservoir and irrigation water isimportant for preservation of water resources.
There are various methods to calculate an amount of
evaporation such as Thornthwaite method (Thornthwaite,
1948) and the Penman method (Penman, 1948), etc.
The relation between evaporation and isotope is discussed[1], [2]. The estimation method of evaporation rate using
isotopic ratio was made by Allison et al. [3] or Gibson et al.
[4]. However the method included many parameters such as
humidity, partition coefficient, oxygen isotopic ratio ofvapor, kinetic isotope effect, resistance of diffusion, and soon shown in expression (1) [5], [6] and then the someparameter is very difficult to measure or estimate.
Therefore, estimation method of evaporation rate tocalculate from isotope is very difficult because evaporation
rate is controlled by many parameters.

δL : isotopic ratio of waterf : ratio of water that remains for the first waterδA : isotopic ratio of vapor
hA : humidity
a : water activity

However, if there is some effective parameter todetermine evaporation rate using isotope, we can estimate
evaporation rate from limited parameter to measure easy.
Therefore, the purpose of this study is to develop simple
method for estimating evaporation rates of water using
oxygen isotopic ratio and finding effective parameter for
calculation of evaporation rate [7]–[9]. Then, we canestimate the oxygen isotopic ratio of residual water whenthe evaporation rate of water reaches 1%.
2. EXPERIMENT AND ANALYSIS

1. INTRODUCTIONd δLh A δL δA  / 1 δL    *a  h A a   vap liq  ei , L / e  ei / e d ln fε* : aε +⊿ε (ε：equilibrium vapor pressure ratio，⊿ε：amount of change equilibrium vapor pressure ratio)
α vap – liq : partition coefficiente i ,L : diffusion resistance of heavy water in watere : diffusion resistance of steam in atmospheric (Constant)e i : diffusion resistance of heavy water in atmospheric(1)

In laboratory test, amount of average evaporation of day,temperature, humidity, and saturation deficit were
measured and then the relation of oxygen isotopic ratio ofwater and their parameters was analyzed.
Table-1 shows the environmental condition and the actual
measurement value of the test.
The 33 evaporation tests were performed between 2005
and 2010 in the drying chamber with uniform temperature,
the refrigerator with low temperature and the rooftop of thebuilding in the Wakayama University. Figure-1 shows the
pictures of evaporation experiment. The water used for theexperiment was tap water of Wakayama University.
Both of the 5L (170 mm in the diameter, 300mm in theheight, and 75 mm in the mouth diameter) and the 2L (126mm in the diameter, 245 mm in the height, and 75 mm ofmouth in the diameter) bottles were used for checking
influence of difference of water volume on the relation
between isotopic ratio of water and evaporation rate underthe experiment.
Figure-2 shows the flow of the experiment. The quantityof water is measured at the intervals of several days during
evaporation period and then 20 ml water was sampled fromthe residual water for clarify relation between evaporation
rate and the oxygen isotopic ratio.
After the carbon dioxide equilibrium method, the oxygenisotopic ratio of the sampled water was measured with mass
spectrometer (Finnigan Mat Delta Plus). The measurement
error of oxygen isotopic ratio (δ18O) is ±0.1‰. The
temperature and humidity were every hour measured andthe data was memorized with the storage meter (Sato meterfactory SK-L200TH). Moreover, the temperature and thehumidity of the outdoor experiment period were used the457

Tab.1 Environmental condition and actual measurement value of the evaporation experiment
Vessel

1
2
3
4
5
6
7
8
9
10
11
12
13
14
15
16
17
18
19
20
21
22
23
24
25
26
27
28
29
30
31
32
33

Evaporation
Place

Average
Average Humidity Saturation Deficit
Temperature (℃)(%)(g/m3)

indoor5L

outdoorindoor2L
outdoorRefrigerator

8.8
9.9
19.3
23.8
24.8
24.9
29.9
24.6
24.9
10.4
7.8
13.1
24.7
27.2
23.9
7.1
24.7
13.0
9.0
19.8
24.8
24.9
24.9
24.7
25.0
7.1
27.6
21.7
7.1
17.2
17.2
4.6
5.5

68.7
65.3
74.3
74.6
47.3
42.3
41.1
68.0
39.8
73.4
60.0
62.1
66.5
67.5
65.5
59.9
66.5
61.9
69.0
72.8
74.8
53.8
42.1
59.2
42.6
59.9
67.6
63.5
59.9
64.5
64.5
25.3
89.6

【drying chamber】

Amount of average
evaporation ofday(mm/day)Start Date

End Date

0.8
0.7
1.4
1.4
3.6
3.5
3.5
2.0
3.3
0.6
0.7
1.0
0.7
1.5
2.4
0.6
0.7
0.9
0.8
1.4
1.4
5.5
5.7
3.6
4.2
0.8
1.9
2.2
1.0
2.3
2.6
0.4
0.2

2007/11/8
2005/12/15
2008/4/21
2008/7/11
2008/10/22
2008/12/15
2005/11/3
2009/4/25
2009/11/19
2006/7/4
2007/11/22
2006/10/24
2007/6/11
2008/7/1
2009/4/30
2007/11/22
2007/6/11
2006/10/24
2007/11/8
2008/4/21
2008/7/11
2008/10/22
2008/12/15
2009/4/25
2009/11/19
2007/11/22
2008/7/1
2009/4/30
2007/11/22
2009/10/9
2009/10/9
2009/7/24
2008/7/14

2008/4/11
2006/4/28
2008/7/9
2008/9/17
2008/12/1
2009/1/23
2005/12/5
2009/8/5
2010/1/28
2007/1/4
2008/4/8
2007/1/4
2007/10/29
2008/9/17
2009/8/27
2008/3/11
2007/10/29
2007/1/4
2008/2/27
2008/6/23
2008/9/17
2008/11/18
2009/1/22
2009/6/5
2009/12/21
2008/3/11
2008/8/29
2009/8/3
2008/3/11
2009/11/9
2009/11/9
2010/3/8
2009/1/20

2.3
2.7
3.6
4.7
11.2
12.5
17.2
6.3
13.1
2.1
2.8
3.7
6.7
7.5
6.6
2.6
6.7
3.7
2.3
3.9
4.9
9.7
12.6
8.3
12.5
2.6
7.6
6.1
2.6
4.5
4.5
4.6
0.6

【refrigerator】【rooftop in the Wakayama University】

Fig.1 pictures of evaporation experimentmeteorological data ofMeteorological Agency.

evaporation

Wakayama

City

of

the

…

Weight measure

Water sampling
Analysis δ18O

Fig.2 Flow of the evaporation experiment

3. RESULT
3.1 Factor in which it affect oxygen isotopic ratio
Figure-3 shows relationship between the amount ofchange of the δ18O per unit evaporation rate and eachamount of average evaporation of day, saturation deficit,
average temperature and average humidity.

458

5L

2L

5L

the amount of change of the δ18O
per unit evaporation rate(‰/%)

the amount of change of the δ18O
per unit evaporation rate( ‰/%)

0.35
0.30
0.25
0.20
0.15
0.10
0.05

0.30
0.25
0.20
0.15
0.10
0.05
0.00

0.00
0.0

5.0

10.0

15.0

0.0

20.0

Indoor

1.0

2.0

3.0

4.0

5.0

6.0

amount of average evaporation of day (mm/day)saturation deficit(g/m3)

IndoorOutdoorthe amount of change of the δ18O
per unit evaporation rate(‰/%)

0.35
the amount of change of the δ18O
per unit evaporation rate(‰/%)

2L

0.35

0.30
0.25
0.20
0.15
0.10
0.05
0.00

Outdoor

0.35
0.30
0.25
0.20
0.15
0.10
0.05
0.00

0

5

10

15

20

25

30

0

temperature(℃)

20

40

60

80

100

humidity(％)

Fig.3 relationship between the amount of change of the δ18O per unit evaporation rate and each amount of
average evaporation of day, saturation deficit, average temperature and average humidity

From the test, negative correlation (R= -0.84) between the
average humidity and the amount of change of the δ18O per
unit evaporation rate was found. In particular, the amountof change of the δ18O per unit evaporation rate is sensitive
to humidity under the low humidity condition and thenchange of the δ18O per unit evaporation rate of residual
water is large during evaporation process under a dry area.
The results of the outdoor experiments were not greatly
different from the indoor laboratory results.
Big differences between outdoor and indoor conditions
were wind and air volume because temperature andhumidity were controlled or measured under the bothindoor and outdoor experiments. Therefore, humidity
condition was more important than wind and air volume
condition.

seasonal change was clarified to affect apparently theamount of change of the δ18O per unit evaporation rate.
Figure-5 shows relationship of the amount of change ofthe δ18O per unit the amount of change of evaporation rateand humidity. The amount of change of the δ18O per unit
the amount of change of evaporation rate changed from 0.2
to -0.2 ‰ / % with evaporation rate under the condition of
more than 60 % humidity. And then, the amount of changeof the δ18O per unit the amount of change of evaporation
rate changed greatly as humidity increased.
On the other hand, under the condition of less than 60 %humidity, the amount of change of the δ18O per unit theamount of change of evaporation rate kept uniform
although evaporation rate changed and then it was inagreement with the equation (1).

3.2 Variation of the amount of change of the δ18O per
unit evaporation rate
Figure-4 shows the monthly variation of the amount ofchange of the δ18O per unit evaporation in summer (2008.4～2008.7) and in winter(2008.12～2009.1). The humidity
was high in summer and low in winter and then, the amountof change of the δ18O per unit evaporation rate increased
from -0.1 to 0.3 ‰ / % as humidity decreased from 80 to
40 %. Therefore, the humidity change accompanying

3.3 Verification of the measurement method in the
Inawashiro lake in Fukushima
The amount of evaporation per year in the Inawashiro
Lake in Fukushima was estimated from the relationship of
the amount of change of the δ18O per unit the amount ofchange of evaporation rate and humidity.
Figure-6 shows distribution of δ18O of the water in the
Inawashiro Lake and rivers surrounding the lake. The area

459

0.30

60

0.20

40

0.10

20

0.00

0

May

Jun

Jul

2008.

experiment6
experiment23
monthly mean humidity(%)

0.50

100

60

0.20

40

0.10

20

0.00

0

0.00
‐0.10

0

0.30
0.20
0.10
0.00
‐0.10

0

- 9.2

-9.1

-9.1
-9.1

HarukawaRiver
原川

Nagasegawa
River

北側河川

-9.2

-9.1
-9.1

-9.1
-9.2
-9.2

-10.1
-10.1

20

40

‐0.20

60

80

-9.1

humidity(%)

60～70%

the amount of change of the δ18O
per unit evaporation rate(‰/%)

70～80%

80～90%

0.40
0.30
0.20
0.10
0.00
‐0.10

0

20

‐0.20

40

60

80

100

humidity(%)

Fig.5 relationship of the amount of change ofδ18O per unit the amount of change of
evaporation rate and humidity

--9.3
9.3
安積疏水
Asaka Canal
Maekawa
前川 -9.8-9.8
River

area: 103 km2
pondage: 3.9 billion m3
amount of the inflow during the year to the lake: 10300 million m3 / year
detention period: 3.7year
average humidity: 74%
δ18O of lake before evaporation: -10.3‰
δ18O of lake after evaporation: -9.3‰

Atagogawa

愛宕川
River

南側河川

Tokonatsugawa
-9.5
-9.5
仲川 舟津川River常夏川
菅川 Nakagawa
Sugegawa
Funatsugawa-9.9 River -9.4
-9.7 River-9.9-9.8 River

-9.8

100

-10.6
- 10.6

小黒川
River

長瀬川

-9.2

100

evaporation rate

is 103 km . Pondage is 3.9 billion m . Maximum depth isabout 94 m. The surrounding length is 55 km. It is amount10300 million m3 / year of the inflow during the year to thelake. The detention period is estimated by dividing theamount of reservoir during year by the amount of the inflow,
and the estimated value is about 3.7 years.
As the δ18O value of the lake was -9.3 ‰ and the δ18O
value of the inflow rivers was -10.3 ‰, the δ18O of water
increased from -10.3 ‰ to -9.3 ‰ by evaporation process[8].

Koishigahama
water
gate
小石ヶ浜水門

80

0.40

ＤＣ

3

-10.5
- 10.5-10.5- 10.5 Kokurogawa

60

evaporation rate
30～40% 40～50% 50～60%-40

Fig.4 monthly variation of the amount of changeof the δ18O per unit evaporation

Takahashigawa
高橋川
River

40

humidity(%)

0.50

2

20

-20

2008.

-0.20

0.10

0.50

0.30

Jan

0.20

-40

80

Dec

0.30

‐0.20

-20

0.40

-0.10

0.40

ＤＣ

the amount of change of the δ18O
per unit evaporation rate(‰/%)

Apr

the amount of change of the δ18O
per unit evaporation rate(‰/%)

80

-0.20

the amount of change of the δ18O
per unit evaporation(‰/%)

100

0.40

-0.10

evaporation rate
10～20% 20～30%

0.50

monthly mean humidity(%)

the amount of change of the δ18O
per unit evaporation(‰/%)

0.50

0～10%

monthly mean humidity(％)

experiment3
experiment20
monthly mean humidity(%)-9.4

Fig.6 distribution of δ18O of the water in the Inawashiro
Lake and rivers surrounding the lake

-9.7

（単位 ‰）

460

humidity (%)

the amount of change of the δ18O per unit evaporation rate (‰/%)

evaporation rate (%) = the amount of change of the δ18O(‰)/ the amount of change of the δ18O per unit evaporation rate (‰/%))
amount of evaporation (mm/year) = the amount of the inflow during the year to lake (km3/year)
× evaporation rate (%) / the area of lake(km2)
Fig.7 flow of the calculation of the amount of evaporation

The amount of change of the δ18O evaporation process
was 1 ‰.
Also the amount of evaporation of the lake was calculated
from the method of Jacobs (1975) with the aerodynamical
technique and the estimated value was 600 mm/year [10].
In the Inawashiro Lake, the average humidity for a year
was 74 %. And then, the amount of change of the δ18O per
unit evaporation rate with 74 % humidity was adopted.
Figure-7 shows a flow of the calculation of the amount of
evaporation.
1) The amount of change of the δ18O per unit evaporation
rate is determined from the figure-5 and average local
humidity.
2) The evaporation rate is calculated from the amount ofchange of the δ18O and the amount of change of theδ18O per unit evaporation rate (‰ / %)3) The amount of evaporation for a year was calculated
from the amount of the inflow during the year tolake, evaporation rate, and the area of lake.
The amount of change of the δ18O per unit the amount ofchange of evaporation rate was calculated from relationshipbetween the amount of change of δ18O per unit the amountof change of evaporation rate and humidity. And then, theamount of change of the δ18O per unit evaporation rate with
74 % humidity, average humidty in the Inawashiro Lake
was 0.15 ‰ / % under the evaporation rate 0～10 %.
The evaporation rate was calculated from the amount ofchange of the δ18O and the amount of change of the δ18O per
unit evaporation rate. Therefore, evaporation rate in the
Inawashiro Lake was calculated to be 6.7 % from 1 ‰, theamount of change of the δ18O and 0.15 ‰ / % , the amountof change of the δ18O per unit evaporation rate shown in
expression (2).
1 (‰) / 0.15 (‰ / %) ≒ 6.7 (%)(2)

The amount of evaporation for a year was calculated fromthe amount of the inflow during the year to lake,
evaporation rate, and the area of lake. Therefore, theamount of evaporation for a year in the Inawashiro Lake

was calculated to be 670 mm / year from 10300 million m3/
year, the amount of the inflow during the year to lake and6.7 %, evaporation rate and 103 km2, the area of the lakeshown in expression (3).
10300 million (m3 / year) × 6.7 (%)/ 103 (km2) = 670 (mm / year)(3)

The estimated value was about 670 mm / year inagreement with the calculation result of the previous
research using the method of Jacobs, 600 mm / year.
However when the evaporation rate in the lake exceed
10%, evaporation rate in the lake can be not calculatedusing only 0.15 ‰ / % because of the amount of change ofthe δ18O per unit the amount of change of evaporation ratechanged with evaporation rate under the condition of morethan 60 % humidity. Therefore, if evaporation rate in thelake exceed 10 %, the amount of change of the δ18O per unit
the amount of change of evaporation rate needs to use
depending on the evaporation rate.
4 CONCLUSION

The purpose of this study is to develop simple method for
estimating evaporation rates of water using oxygen isotopic
ratio. From the empirical results, negative correlation (R=-0.84) between the average humidity and the amount ofchange of the δ18O per rate of unit evaporation was clarified.
And the amount of change of the δ18O per unit the amountof change of evaporation rate changed with evaporation rate
under the condition of more than 60 % humidity.
The amount of evaporation per year in the Inawashiro lakein Fukushima was estimated to be about 670 mm / year
from the relation between humid and evaporation rate andthe estimated value was in agreement with the calculation
result of the previous research using the method of Jacobs,
600 mm / year.
Therefore, the simple estimation method using only
oxygen isotopic ratio and humidity is effective for
estimation of evaporation rate of an actual lake or a pond.
And when evaporation rate in the lake exceed 10 %, we
can estimate the evaporation rate from the amount ofchange of the δ18O per unit the amount of change of
evaporation rate changed with evaporation rate.

461

5. REFERENCES[1]

Machida I, Kondoh A: Stable Isotope Ratios of Natural Water inJapan –The Analysis by Using Environmental Isotopes Database, J.
Japan Soc. Hydrol.＆ Water Resour. Vol.16，No.15, pp556-569,
2003.

[2]

Yoshimura K, Fujita K, Kurita N, Abe O: International Workshop
on Isotopic Effects in Evaporation, -Revisiting the Craig-Gordon
Model Four Decades after its Formulation - Participation report,
Journal of Japan Society of Hydrology & Water Resources.19,
pp.420-423, 2006.

[3]

Allison, G.B., Brown, R.M. and Fritz, P : Estimation of the isotopic
composition of lake evaporate., Jour, of Hydrol. , 42, pp.109-127,
1979.

[4]

Gibson, J.J., Edwards, T.W.D., Bursey, G.G. and Prowse, T.D. :Estimating evaporation using stable isotopes: quantitative results andsensitivity analysis for two catchments in northern Canada, 24,
pp.79-94, 1993.

[5]

Fontes, J. –ch. and Gonfiantini, R.: Comportment isotopique ancours de levaporation de deux

basins Sahariens, Earth Plan.Sci.

Letters, 3, pp.277-374, 1967.
[6]

Pupezin, J., Jancso, G. and Van Hook, W.A.: Vapor pressure isotope
effect in aqueous systems. I, Jour. Phys. hem. , 76, pp.743-762.

[7]

Miyaji K, Okada S, Ii H, Nagabayashi H: Proposal of simple
measurement method for evaporation rate by using oxygen isotopic
ratio in the Inawashiro Lake., Annual Journal of HydraulicEngineering., Vol.52, pp.289-294, 2008.

[8]

Miyaji K, Ii H, Miyahara S: Proposal of method for estimating theevaporation rate in lake by using oxygen isotopic ratios andhumidity. , Annual Journal of Hydraulic Engineering., Vol.53,
pp.331-336, 2009.

[9]

Miyahara S, Ii H: Proposal method for estimating evaporation rates
of water using characteristic of change of oxygen isotopic ratio,
Japan Geoscience Union Meeting 2011.

[10] Watanabe A: The role of heat of the Inawashiro Lake, Tenki, 30 (3),pp.137-142, 1983.

462

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Landslide Susceptibility Mapping using Logistic Regression Model With
Neighborhood Analysis: A Case Study in Mizunami City
Wang Liangjie. Sawada Kazuhide. Moriguchi Shuji., Department of Civil Engineering, Gifu University, Yanagido
1-1, Gifu, 501-1193, JapanABSTRACT: Landslides which affect human lives and economic
losses are always attracted a lot of concerning in modern society. In
order to identify the potential hazardous areas related to landslides,
three methods have been used, such as qualitative or
knowledge-based
method,deterministic
methodandquantitative-based method. Geographical information system (GIS)
technology and high computing ability provide a convenient tool todeal with landslide triggering factors and make thequantitative-based method achieve effectively.
In this study, Landslide-related factors such as topographicalelevation, slope angle, slope aspect, topographical wetness index(TWI) and stream power index (SPI), were employed in thelandslide susceptibility analysis. The logistical regression was used
to obtain the relationships for landslide susceptibility between
landslides and causative factors. The distributions of observed
landslides were used to evaluate the performance of thesusceptibility map. The approaches described in this paper showed
us that the logistical regression and neighborhood can be used as
simple tools to predict the potential landslide locations. And thismap will be helpful for city planning, infrastructure construction andagriculture developments in the future.

In this study, neighborhood analysis “seed cell approach”
proposed by (Suzen and Doyuran., 2004) [4] and logistic
model were applied to create the relationship between
landslides and controlling factors with GIS in Mizunami
city ,Gifu prefecture where is well know for landslide
hazards.

Keywords: Landslide, Susceptibility map, Logistic regression,
GIS

2.1 Landslide Related Factors
The first database is topographic factors. We create digital
elevation model (DEM) derive from a triangulated irregular
network (TIN) using elevation points from GSI (Geological
Survey of Japan AIST) with spatial resolution 10m. The
parameters such as elevation, slope angle, aspect, plancurvature, profile curvature were constructed from DEM. The
second database is water-related factors. Those parameters
were also obtained from DEM. The topographical wetness
index (TWI) and stream power index (SPI) are importantparameters of wetness and stream power [5] (Moore et al.,1991) and these are defined as following formulas.

1. INTRODUCTION
Landslides which cause the loss of human life and the damage
to the social economy were attracted a lot of attention over thelast decades. According a previous study Schuster (1996) [1],landslides will increase in the next decades due to continued
deforesting and the changing climatic patterns inlandslide-prone areas. In order to identify thelandslide-related area, there are three main approaches toassess the landslide susceptibility: qualitative methods,
deterministic methods and quantitative methods. In the late1970s, qualitative approaches were widely applied byengineering geologists to evaluate landslide. Deterministic
approaches focus on slope geometry, shear strength data, andpore-water related data (Netra et al., 2010) [2] but lack oftaking climatic and human induced factors into accounted.
Nowadays, the rapid developments of computer technology
and geographic information system (GIS) provide aconvenient tool to deal with landslide triggering factors andmake the quantitative-based method achieve effectively.
Among a lot of quantitative methods, logistic model wasrecognized as the suitable approach to assess landslide
susceptibility because it is free of data distribution and canhandle a variety of datasets (Nandi et al., 2009) [3].

2. DATABASE CONSTRUCTION
In order to detect the landslide-related factors, some priorknowledge and experiments should be prepared. The main
factors include three aspects such as topographic factors
water-related factors and human being activity factor are
applied in this study. The approach mentioned is based on theassumption that past environmental conditions at the time oflandslides can be keys to evaluation the potential sites forlandslide in the future. Then, all the landslide-related factors
were created and store in the spatial database by using a GIS
software with the pixel size or mesh 10×10m.

TWI  loge ( A / b tan  )(1)

SPI  A tan  / b(2)

Where A (m2) is the upstream area, b is the size of the meshexpressed as m. and α is the slope gradient. The third databaseis the human actively factors. The proximate to the highway
was used to indicate the human actives induce the landslides.
The table 1 shows the significant of the all parameters related
to landslides and their spatial distribution patterns are shownin Fig.1.
2.2 Landslide Inventory and Neighborhood Analysis
The landslide inventory map of the study area wasdownloaded from the organization of NIED (National

463

Table 1 Landslide causative factors
Data type

Topographic

Water-related

Anthropogenic

Factors
Elevation
Slope
AspectPlan curvature
Profile curvature
Tangent curvature
Solar radiation
Flow length
Flow direction
Flow accumulation
Stream power index
Distance to rivers
Topographic wetness index
Distance to highway

SignificanceClimate, Vegetation
sub-surface flow velocity
Evapo-transpiration
Converging, soil water content ,soil characteristics
Erosion ,Geomorphology
Erosion and deposition
Weathering soil moisture
Run off velocity, run off volume
Runoff velocity. potential energy
Runoff velocity and potential emerge
Erosive power of water flow
Susceptible to hill slope undercutting
Soil water content
Landslide triggering by road construction and vibration

Fig.1. Distribution of landslide causative factors

464

Research Institute for Earth Science and Disaster Prevention).
The sizes of landslides rage from 0.00002 km2 to 1.2 km2,
which cover 13.6% of the study area. According to Varnes’s(1978) definition, mass movement like soil slides, debris
slides, rock slides and debris flows are incorporated into theterm landslides.
Neighborhood analysis is a spatial analysis tool used to
identify the “seed cells” which represent the undestroyedmorphologic conditions before the landslides occurred(Suzen and Doyuran., 2004) [4]. These “seed cells” can beextracted from the boundaries of landslides by buffering zone
to the landslides. Former studies show us that the buffer zones
should be range from 100m to 150m. In this study, 120m
buffer zone is chose to be the “seed cells”.

statistic significance if there are no validate points used to
identify the efficient of the result. Therefore, the 3518
landslide points occurred in the past were employed tovalidate the accuracy of the landslide hazard index map. Fig.2
show us the distribution of landslide points, “seed cells” andlandslide free points in Mizunami city.

3 LANDSLIDE SUSCETIBILITY MAPPING
3.1 Logistic Regression Model
Logistic regression model is applied to establish therelationship between a dependent variable and independent
variables (Atkinsm and Massari., 1998) [6]. This model iscareless about the distribution pattern of the independent
variables. Most of the topographic factors don’t have normal
distributions. Therefore, logistic regression model could get
better results comparing other mathematic models. The
predicted values range from 0 to 1 can be defined as landslide
hazard index. The index can be expressed as following
formula.

P  1 /(1  e  z )(3)z  b0  b1 x1  b2 x2  ...  bn xn(4)

Fig.2. Distribution of sampling points

Where P is the probability of landslide occurrence (landslide
hazard index), z is the linear logistic mode, b0 is the interceptof the model, n is the number of landslide causative factors, bn
is the weight of the each factor and xn is the landslidecausative factors.
Our purpose of this research is to propose a model that could
be applied in some place where people could not easily getaccess to or some cities which lack of basic datasets (landcover dataset or digital geological dataset). In this study, mostof the landslide-relative factors derive form the DEM that
could be easier get from nowadays spatial technology. The
causative factors are continuous that would avoid theoccurrence of huge redundant data.
3.2 Landslide Points Sampling
Former studies show us that it is the best sampling pattern if
the ratio of landslide points to landslide free points is equal to1(Suzen and Doyuran.,2004) [4]. For this purpose, the ratio is
equal to 1 between “seed cells” points that used to extract the
attributes from the landslide-causative factors and the random
landslide-free points. All of the 11266 points were chosen to
create the relationship between landslides and causative
factors in logistic regression model. The model would lose his

3.3 Model Results and Validation
The forward stepwise logistic regression model was appliedin this study. The probability for stepwise was range from0.05 to 0.1. And then, χ2 value of the Hosmer and Lemeshow
Test, Cox & Snell R Square and Nagelkerke R Square wereapplied to evaluate the effectiveness of training datasets. The
table 2 shows that the result of the model had statistic
significance and the independent variables could explain thedependent variables in this model. During the process ofmodel calculation, the Wald value decreed from 565.8 to 5.3
in all 8 steps. The distance to highway which had the highest
value of Wald value implied that this landslide-related factor
was the most significant value for explaining landslide
occurred, followed by distance to rivers, flow length,
elevation, slope angle, solar radiation, and flow accumulation.
The relationship between the causative factors and landslide
could be shown in table 3. According to table 3, slope angle,
flow length and solar radiation were positive relationshipwith the occurrence of a landslide whereas flow accumulation,
distance to highway, elevation and distance to rivers indicatednegative relate to the occurrence of a landslide.
Using the weights shown in table 3, the landslide hazard
index (LHI) of landslides was calculated (Fig.3).

465

Table 2 Statistics of logistic regression model with landslide causative factors
Hosmer and Lemeshow Test
χ2
Sig.
Sig.
173.6648.000-2 Log likelihood

Cox & Snell R2Nagelkerke R214598.819

0.286

0.315

Table 3 Coefficients of the logistic regression model for the seven landslide-related factors
Coefficient

Independent variables
Slope angle

0.01877

Flow length

0.00006

Flow accumulation-0.000003

Solar radiation

0.000002

Distance to highway-0.00007

elevation-0.00289

Distance to rivers-0.00044

Constant-0.19102

Fig.3. Landslide hazard index map

Fig.4. Reclassified landslide susceptibility map

466

4 CONCLUSION
The landslide hazard index decreased from 0.94 to 0.06. If thevalue is close to 1, it indicates landslides are more likely
happen in this area.
The accuracy of the logistic regression model of landslide
susceptibility was evaluated by calculating relative operative
characteristic (ROC) and percentage of observed landslides
points in various susceptibility categories [3] (A.Nandi andA.shakoor., 2009). The area under ROC cure (AUC) value
which represents the quality of the probabilistic model bydescribing its ability to predict the occurrence or
non-occurrence of an event was applied in this paper(Yesilnaca and Topal., 2005) [7]. The value of AUC is range
from 0.5 to 1. If the AUC value close to 1, that means highaccuracy of prediction model and if the AUC value close to
0.5 which indicates the inaccuracy of the model (Fawccett.
2006) [8]. In this study, we obtained the AUC value was 0.69,
which implied the reliable correlation between causative
factors and landslides. Also the observed landslide points
were used to identify the accuracy of the landslide probility
map. We divided the landslide hazard index map into fivechasses: very low (0.05<LHI≤0.1), low (0.1<LHI≤0.3),
medium (0.3<LHI≤0.5), high (0.5<LHI≤0.7) and very high
(LHI≥0.7).
From fig.5, 9.55% of observed landslide points were found invery high susceptibility class. 6.76% of the landslide points
occur in the areas with very low and low susceptibility classes.
Near 60% of the landslide points were concentrated in thearea with high susceptibility class. The hypothesis of themethod was that a landslide would occur in area with at least
medium values of susceptibility or high and very highsusceptibility values (Bai et al., 2010) [9]. 93.2% of landslide
points appeared in the area with susceptibility classes frommedium to very high. Among the whole Mizunami city, 12%of the study area was designated as very high susceptibility
zone which had the approximately equal percentage of theformer landslide area (covering 13% of the study area)
indicated that the dividing approach was appropriate in
Mizunami city.

Landslides are very complexity phenomenon in the natureworld. The reasons cause landslides appearance are hardly beunderstood. Although some landslide causative factors could
be obtained by exploiting advanced spatial technology, somefactors remain unknown. There are many quantitative
methods used to create the relationship between landslides
and causative factors. And then the landslide susceptibility
maps are applied for mitigating the geo-hazard in the world.
In this paper, logistic regression model which is simple andcareless about the distribution pattern of the independent
variables was closed to build the landslide susceptibility map
by using GIS tools. Most of the causative factors were derived
from DEM which was easier acquired nowadays. Seed cells
which extracted from the boundaries of landslides were
employed to represent undestroyed geomorphologic
conditions. The seed cells points were entered into the logistic
regression model as training datasets and the observed
landslides points were applied as the test dataset to evaluate
the accuracy of the model. The results of landslide
susceptibility analysis showed us that 70% of the observed
landslides points were located in high and very highsusceptibility categories. The model seems to be reliable in
Mizunami city. The accuracy and prediction ability of thelandslide susceptibility map could offer us crucial
information for city planning, infrastructure construction andagriculture developments in the future or in other area withsimilar conditions.
5 REFERENCES
[1]

[2]
[3]
[4]

Landslide Frequency (%)

59.81

60

[5]
50

[6]

40

30

[7]

23.88

20
9.55

10

6.59

[9]

0.17

0
Very Low[8]

Low

MediumHigh

Very HighLandslide Susceptibility
Fig.5. percentages of observed landslide points
falling into different susceptibility categories
467

Schuster, R., Socioeconomic significance of landslides. In: Turner,
A.K., Schuster, R.L. (Eds.), Landslides: Investigation and Mitigation,
Special Report, vol. 247. National Academic Press, Washington, DC,
1996, pp. 12–36.
Netra, R. R., John, R.G., John, D.V., Assessing susceptibility tolandslide: Using models to understand observed changes in slopes.
Geomorphology, 2010, 122 (4), 25–38.
Nandi, A., Shakoor, A., A GIS-based landslide susceptibility
evaluation using bivariate and multivariate statistical analyses.
Engineering Geology, 2009, 110, 11–20.
Süzen, M.L., Doyuran, V., Data driven bivariate landslide
susceptibility assessment using Geographical Information Systems: amethod and application to Asarsuyu catchment, Turkey. Engineering
Geology, 2004, 71, 303–321.
Moore, I.D., Grayson, R.B., Ladson, A.R., Digital terrain modeling — areview of hydrological, geomorphological, and biological applications.
Hydrological Processes, 1991, 5, 3–30.
Atkinson, P.M., Massari, R., Generalized linear modeling ofsusceptibility to landsliding in the central Apennines, Italy. Computers
and Geosciences, 1998, 24, 373–385.
Yesilnacar, E., Topal, T., Landslide susceptibility mapping: acomparison of logistic regression and neural networks methods in a
medium scale study, Hendek region (Turkey). Eng. Geol. 2005, 79, 251–266.
Fawcett, T., An introduction to ROC analysis. Pattern Recognition
Letters, 2006, 27, 861–874.
Bai, S.B., Wang, J., Lü, G.N., Zhou, P.G., Hou, S.S., Xu, S.N.,
GIS-based logistic regression for landslide-susceptibility mapping ofthe Zhongxian segment in the Three Gorges Area, China.
Geomorphology, 2010, 115, 23–31.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

The Evaluation of the River Environment and Water Quality by Image
Analysis in the Yamato River Basin in Japan
Masanobu TANIGUCHI, Hiroyuki II, Tatemasa HIRATA
Department of Environmental Systems, Faculty of Systems Engineering, Wakayama University



ABSTRACT: Amounts of attached algae and the river bed
area were estimated by image analysis in the Yamato River
Basin and then natural cleaning effect of nitrogen in the river
was evaluated. The image analyze method has beendeveloped since 2009. The Yamato River Basin has 1070 km2
areas and it is difficult to understand the distribution of plant,sand place, depth and attached algae. The 75% BOD value(Biochemical Oxygen Demand) annually in the YamatoRiver was the worst of all Japanese rivers managed by theJapanese government. BOD values, ammonia nitrogen andanion surfactant concentrations were high in winter and low
in summer and they decreased with temperature. Therefore, it
is important to understand amount of natural cleaning effectand the cleaning place. As a result, many adhesion algae lives
in less than 30cm depth, the amount of adhesion seaweed was
calculated by divided color by based color plus-minus 20%
each RGB color code. So, the rate of decrease of nitrogen was
about 0.4 mg/l per km in summer, it occurred in less than
30cm depth.

2004) Therefore, the origin of the river water and thegroundwater was analyzed using environmental isotopes andthe chemical composition of the Ishikawa River of the
Yamato River basin, and the character of the water quality
was investigated using field survey data from 1999 to 2005 in
the Yamato River. (Hiroyuki II et al, 1999) Moreover, adecrease in water temperature caused an increase in BOD
values, ammonia nitrogen and anion surfactant
concentrations because these concentrations were high in
winter and were low in summer (Masanobu Taniguchi et al,
2002 and 2005). In this research, we developed the image
analysis software, and evaluated the river environment andwater quality by the image analysis.
2. STUDY AREA AND METHOD

Keywords: evaluation, water quality, image, the YamatoRiver
1. INTRODUCTION
The Yamato River was mainly polluted by sewage water andindustrial effluent and has a major water pollution problem inJapan. The catchment consists of 37% mixture of coarse
fragment, sand and clay, 36% granitic rock, 8% alternate
layer. 2,000,000 people live in the Yamato River catchment.
The amount of water demand in the catchment was larger
than the amount of effective water supply, and then theamount of shortage water was provided from the neighboringcatchments. The cause of the wastewater pollution was
overdue development of sewage systems in recent decades.
The 75% BOD value (Biochemical Oxygen Demand)
annually in the Yamato River was the worst of all Japaneserivers managed by the Japanese government. Because of this,governments carried out a cleaning campaign to improve thewater quality in the Yamato River from 1985 to
2005.(Yamato River construction office, 2005) However the
75% BOD value was still bad and exceeded the allowable
limits set by the Japanese environmental quality
standard.( Ministry of Land, Infrastructure and Transport,

Fig. 1 the geological information of study area and observing
site
Figure 1 shows the geological information of study area andobserving site. The surround red area is mountain area
composed of granite. The Nara basin consists of mad, sandand stone. The river bed of the Yamato River Basin consistsof weathering granite. Where river bed is sand, the river water
is easy to permeate and then plants and algae or other microlife grow. In this study, the distribution of sand area isimportant, and rate of sand area and living algae area isestimated by image analysis.
Long term concentrations were analyzed using data from thefixed observing station of the National Land andTransportation Ministry from 2006 to 2009 and data of a fieldsurvey by Wakayama University from 2010 to 2011.

469

Figure 4 shows a time series of T-N, NO3--N and NH4+-N
concentrations from 2006 to 2009. T-N is high in winter andis low in summer as well as the BOD concentration. And
Nitrate nitrogen concentration increased regularly in winter inthe same way as the BOD concentration. And Ammonia
nitrogen concentration increased regularly in winter in thesame way as the BOD concentration. Ammonia nitrogenconcentration is nothing in summer mostly in every year.
Fig. 2 Survey Method
4. NITROGEN OF FLOW DIRECTION CHANGE(mg/l)
10

NH 4+-NNO 3 -N
15
δ N

(‰)

10

濃度 5

3. WATER QUALITY OF ANNUAL CHANGE

15

δ N

Kamenose

Hujii

④

③

②lower weir

②upper weir

◇BOD (mg/L)□DO(mg/L)

5①0Upper Stream

20
ＤＯ(mg/L)

2004 Feb

Conscentration

The river width and depth is measured at 3 sites in the
middle stream of the Yamato River, the lower stream of the
Tatsuta River and the lower stream of Ishikawa River in 2010.
The algae’s living place is related by the water depth,
distribution of shallow place is important. So, we measure theperpendicular of rivers, because color become dark depend on
depth, it is possible that the distribution of depth is divided bycolor. The algae and sand distribution is made by fixed color
comparing survey photographs and the Aerial photographs.

Lower Stream

102010-01

2009-11

2009-09

2009-072009-05

2009-032009-01

2008-11

2008-09

2008-072008-05

2008-03

2008-01

2007-11

2007-09

2007-072007-05

2007-03

2007-01

2006-11

2006-09

2006-072006-05

2006-03

2006-01

0

Fig. 5 Relation between nitrogen isotope ratio and nitric acidand ammonia nitrogen in February 2004

Fig.3 A time series of BOD and DO concentration
Figure 3 shows a time series of BOD and DO concentrationfrom 2006 to 2009. The graph used data of Ministry of Land,
Infrastructure and Transport. One day intensive measurementof DO is performed in winter and summer. DO concentration
doesn’t decrease in winter but decreases about 4mg/L insummer. BOD values were the maximum in winter and theminimum in summer almost generally. BOD means amountof organic compounds and is defined as amount of usedoxygen for decomposition of organic compound. In the
Yamato River, the domestic sewage flow into the river all theyear round, it is need to consider the influence of
decomposition of organic compound. So reaction of using
oxygen is active by living things in summer, DO decrease insummer.

2010/01

2009/01

2008/01◇T‐N(mg/L)◇NO3 ‐‐N(mg/L)□ NH4 +‐N(mg/L)

2007/01

2006/01

8
7
6
5
4
3
2
1
0

Figure 5 shows the relationship between nitrogen isotopic
ratio for nitrate and ammonia nitrogen in February 2004.
There is a big city on the upper stream of the Tatsuta Riverand domestic sewage directly flows into the river.
Most domestic sewage was only treated in combined-typeprivate sewage treatment systems. Sampling points with a bed
and lateral walls of a river covered with concrete are the point
1 and 2 in the Tatsuta River shown in Fig.5. At these points,
the river was always over 1m in depth. However, the bed andlateral walls at the point 3 and 4 in the lower stream of the
Tatsuta River were not covered with concrete but the bed was
alluvial sediments. In the point 3 and 4, the river had a very
natural form. Thus there were many plants on the river bed
and the river maintained just a few cm to a few tens of cm indepth. The total inorganic nitrogen concentration of all areas
in February was higher than those in July. In February,ammonia nitrogen / inorganic nitrogen (ammonia nitrogen +nitrate nitrogen) ratios at all points were higher than those in
July. Each total inorganic nitrogen concentration in point 1
and 2 areas was found to be higher than those in the point 3
and 4 areas, although each nitrate nitrogen concentration in
point 1 and 2 areas was lower than those in the point 3 and 4
areas.

Fig.4 A time series of T-N, NO3—N and NH4+-N
concentrations
470

15
10
10

15

δ N

5
5



(CH 2O)106 (NH 3 )16 H 3PO 4  84.8NO 3  84.8H  

Lower Stream-

NH4  1.86O2  1.98 HCO3 (2)-(0.0181 0.0024)C5H7 NO2  1.04H2O  0.98NO3  1.88H2CO3

Generally Ammonia changes into nitrate in aerobic
conditions and then finally organic compounds change into
nitrate. These reactions depend on activity of livingorganisms. Temperature is important for the activity of theorganisms. However nitrate nitrogen concentration did notchange with temperature. Nitrate acid was consumed byphotosynthesis or denitrification and the biological activitiesof photosynthesis or denitrification were also dependent on
temperature.

(4)(CH 2 O)106 (NH 3 )16 H 3 PO 4  138O 2

When the photosynthesis reaction and denitrification
reaction occurred, Nitrogen was removed in the water.

Sub Stream

3.5

NO3-N
NH4-N

2.5

Ishikawa River

3.0

2.0
1.5
1.0
0.5

Meiji Bridge

Ishikawa Bridge

0.0

(1)

Nitrification in an aerobic condition is shown by thefollowing equation (2).
-

106 CO 2  16NO 3  16H   H 3 PO 4  122H 2 O 

Kunitoyo Bridge

Figure 6 shows the relationship between nitrogen isotopes,
nitrate and ammonia nitrogen in July 2004. Sample points arefrom the upper stream to the lower stream of the Tatsuta Riverand Yamato River’s main stream. The total inorganic
nitrogen concentration of all areas in February is higher thanthose in July. In February, ammonia nitrogen / inorganic
nitrogen (ammonia nitrogen + nitrate nitrogen) ratios at all
points are higher than those in July. Each total inorganic
nitrogen concentration in the point 1 and 2 areas is found to behigher than those in the point 3 and 4 areas, although each
nitrate nitrogen concentration in the point 1 and 2 areas islower than those in the point 3 and 4 areas. Also, eachammonia nitrogen concentration in the point 1 and 2 areas ishigher than those in the point 3 and 4areas. In both winter andsummer season, δ15N increased down the stream and inparticular, the δ15N change from the point 1 to 4 areas is large.
When plants consumed nitrate nitrogen, the δ15N of nitrate isthought to increase.
Generally, organic compounds in an anaerobic condition
were decomposed to ammonia by bacteria. The reaction
formula is given by the following equation (1).

Photosynthesis reaction is shown by the following equation(4).

Kamenose

Fig. 6 Relation between nitrogen isotope ratio and nitric acidand ammonia nitrogen in July 2004

(CH2O)106(NH3 )16 H3PO4  53CH4  53CO2 16NH3  H3PO4(3)106CO 2  42.4N 2  16NH 3  H 3PO 4  148.4H 2OOhji Town OfficeKamenose

Hujii

④

③

②community②lowe r weirUpper Stream②upper weir0①

0

Tatsuta River

Conscentration

3km

④

5km

1km

Nitric acid is an intermediate substance between
decomposition and photosynthesis and denitrification
process.
Therefore, in high-temperature conditions organic
compounds easily change into nitrate by aerobic bacteria andnitrate is consumed by denitrification or photosynthesis. On
the other hand, in low-temperature conditions, organic
compounds are stable.
Denitrification reaction is shown by the following equation
(3).

(‰ )
20

Nhukatabet aka Bridge+

NH4 - NNO3 -N
15
δ N

2004 July

Nitrogen Concentration(mg/L)(mg/l)
15

Fig.7 Relation between nitrate nitrogen and ammonia
nitrogen in July 2011
Figure 7 shows relation between nitrate nitrogen andammonia nitrogen concentration in July 2011. Both nitrate
nitrogen and ammonia nitrogen concentration in 2011 islower than those in 2004. However, ammonia nitrogen isshown in the lower stream of Tatsuta River, because domestic
sewage flows into the river yet. So, it is thought that organic
compound decomposed to ammonia in the Tatsuta River. The
nitrate nitrogen concentration of Meiji Bridge decreased fromupper stream. In the Yamato River, the ammonia nitrogen isnoting, so it is estimated to occur the nitrification. Therefore,
it is estimated that the nitrate acid is an intermediate
substance through both decomposition and photosynthesis
and denitrification process in the Yamato River of middle

471

stream to lower stream. In order to estimate amount of thenitrogen change process of organic compound of
decomposition to photosynthesis or denitrification, it
considered that relation between chromaticity and water
quality.
5. RERATION BETWEEN CHROMATICITY AND
WATER QUALITY
The river chromaticity changes with amount of floating
material, like as phytoplankton, soil, leftover food and humic
acid etc. Chlorophyll a is a coloring matter of the cell ofphotosynthesizes vegetation. Then amount of phytoplankton
increases with Chlorophyll a.
22

10

18

Tatsuta River

Ishikawa River

8

16
14
12
10
8

6

4

6

Chromaticity

Chlorophyll a (μg/L)

Chromaticity
Chlorophyll a

Sub Stream

20

Figure 9 shows the outline of image analysis. If algae occur
or the states of a river bed differ, it is possible that the
distribution of sand and algae is decided by photograph. Inorder to fix the color of algae and sand, the survey photographand the Aerial photographs are compared. The average ofcolor and the near color of plus-minus 20% are used, because
the photograph includes some error.
Table 1 Result of Image analysis
Site
Sand
Algae

Yamato Riv.(Low)17%
19%

Yamato Riv.(Mid) Tatsuta Riv.
15%
2%
41%
55%

②
7%
43%

④
2%
39%

Table 1 shows result of Image analysis. The used data isanalyzed by Google’s data, in objective rough calculation.
It is unknown picture date, it is trial calculate. In this result,algae living place is estimated shown by blue line area in
Figure 9, 19% of lower stream is algae’s living place. The
middle stream of the Yamato River is 41%, Tatsuta River is55%. Comparing survey photographs and the Aerial
photographs, it is estimated that algae's living place area isaverage 46% of calculated area, the range is 14% to 65%.

2

6. CONCLUSION
Meiji Bridge

Ishikawa Bridge

Kunitoyo Bridge

Kamenose

Ohj i Town Office④

Nhukatabetaka Bridge

4

Fig.8 Relation between chromaticity and chlorophyll a in July
2011
Figure 8 shows relation between chromaticity andchlorophyll a in July 2011. The chromaticity is high in Algae
living place. The chlorophyll a tends to change as same as
chromaticity. The chromaticity and chlorophyll a in the
Tatsuta River is lower than the other upper stream. However
the chromaticity and chlorophyll a in the Nukatabetaka
Bridge is high, the several algae are found in the place. Thereis a river purification facility in the place.

Amounts of attached algae and the river bed area were
estimated by image analysis in the Yamato River Basin andthen natural cleaning effect of nitrogen in the river was
evaluated. The image analyze method has been developedsince 2009. BOD values, ammonia nitrogen and anion
surfactant concentrations were high in winter and low insummer and they decreased with temperature. Therefore, it isimportant to understand amount of natural cleaning effect andthe cleaning place. As a result, many adhesion algae lives inless than 30cm depth, the amount of adhesion algae wascalculated by divided color by close to based color
plus-minus 20% each RGB color code. So, it occurred in lessthan 30cm depth, distribution of algae’s living place isestimated.
7. REFFERENCES
[1]
[2]
[3]

[4]

Fig.9 the outline of image analysis
472

Yamato River construction office, http://www.yamato.kkr.mlit.go.jp/
YKNET/index.html
Ministry of Land, Infrastructure and Transport, Japan, Present
condition of water quality of all Japanese first class rivers in 2004, p82
Masanobu Taniguchi, Hiroyuki Ii, Tatemasa Hirata, Masataka
Nishikawa, Yumi Ogawa, 2005. Decompose process of sewage organic
material using nitrogen isotope and load analyze in Yamato River. The
11th Committee on Technology of River, JSCE, pp35-40(in Japanese)
Masanobu Taniguchi, Hiroyuki Ii, Tatemasa Hirata, Analysis oforganic nitrogen compound in domestic sewage in Yamato River using
a social experimental test, 2010, vol. 30, Part 10, p. 1587–1590,
Stuttgart, April 2010

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Distribution of Oxygen and Hydrogen Stable Isotopic Ratios and Estimationof the Spring Water Origin in the Kushiro Moor
Natsuki Kawabe. Hiroyuki Ii., Graduate school of Wakayama University, Wakayama University


ABSTRACT: The purpose of this study is to clarify groundwater
movement in the Kushiro Moor comparing the oxygen and hydrogenstable isotopic ratios for stream river waters around the Kushiro
Moor with those for spring waters in the Kushiro Moor andestimating the origin of spring waters in there. The δ18O values of
stream water range from -8.0 to -11.5 ‰ relatively high in the east
including Kushiro, north and the south Hokkaido. On other hand, theδ18O values range from -11.5 to -14.0 ‰ relatively low in the central
and the northeast Hokkaido. The springs in the Kushiro Moor weredivided into low δ18O values group and high δ18O values group.
Their origins are found to be divided into at least two areas at thenorth and southeast of the moor.

Kushiro moor

Keywords: oxygen and hydrogen stable isotopic ratios,
Kushiro Moor, spring water, water cycle
1. INTRODUCTION
Hokkaido is a northern Japanese island. Its area accounts formore than 20% of the whole of Japan and about 80,000 km2.
In 1980, the Kushiro Moor, the largest wetland in Japan was
registered in the Ramsar Convention as valuable ecosystems[1]. As the Kushiro Moor has some spring inside, then even in
winter there is no freezing place which plays an importantrole in providing some living place for birds in winter. In
order to preserve the wetlands it is necessary to understandthe water cycle in a wide area including the origin of springwaters.
Although groundwater-dependent rate is low, 2.9 percent in
Hokkaido at present, however it may increase with global
warming and then it is important to estimate groundwaterflow [2].
Isotopic ratios of precipitation vary with region. Therefore,
comparing the isotope ratios of precipitation and groundwater,
the origin of ground water is estimated. Isotopic ratios
measurements of precipitation need water samplingthroughout the year, so it is difficult to take a number ofmeasuring points. Isotope ratio of small watershed stream
water can be regarded as an average value of precipitationfrom the previous study [3].
We can estimate the origin of spring comparing the isotope
ratios of spring water with stream water. Isotopic ratios of thestream water vary with the distance from the coast, altitude,
and prevailing winds. A precipitation isotopic ratio for any
place can be estimated from the altitude and distance from thecoast by the local calculation formula in the Kii Peninsula [3]and the Yakushima Isalnd [4]. Therefore, the purpose of thisstudy is to estimate the origin of spring water in the Kushiro
Moor and to determine the local formula for the Hokkaidoarea.

Fig1.Sampling point

2. SUMMARY OF SAMPLE COLLECTION
The study area is the whole area of Hokkaido including theKushiro Moor. Localities of water samples are shown in Fig.1.
The sampling days were in Jun. 13, 2005 to Jun. 18, 2005, Oct.
7, 2005 to Oct. 8, 2005, Sep. 26, 2006 to Sep. 28, 2006, Apr.
27, 2007 to Apr. 30, 2007, Jul. 9, 2007 to Jul. 13, 2007, Sep. 9,
2008 to Sep. 16, 2008, Sep. 11, 2009 to Sep. 17, 2009, Apr. 28,
2010 to Mar. 4, 2010 and Sep. 5, 2010 to Sep. 12, 2010.
There is the Sarobetsu wilderness and a lot of lowlands in thenorth region. In the northeast region, there is the Saroma Lakethat is biggest brackish lake in Japan. In the east region, there
are the Kushiro Moor and the Masyu Lake and spread tolowland in the whole. There are the volcanic highlands in thecentral region, including the Asahi Peak that is the highest
peak in Hokkaido. In the south region, there is the Tokachi
Plain, and Hidaka Mountains are running around west of thisplain. The west region is composed of peninsula and dottedvolcanic terrain [5].
Above all, the Kushiro Moor is about 36km length from northto south and about 25km wide from east to west and the total
area is 189.2 km2. Fig.2 shows the sampling points around theKushiro Moor. In the Kushiro Moor, many springs can beseen around the Chiruwatsunai River located in thenorth-Kushiro Moor. Therefore, sampling of spring and riverwaters in the Kushiro Moor was performed intensively.
In this study, river waters, spring waters, ground waters, andhot spring waters were collected in the whole Hokkaido.
To prevent sampling water from contamination, stream water
sampling was selected because stream water is thought not to473

Kussyaro Lake

-15.0
0

Akan Lake

-13.0

-11.0

-9.0

-7.0

-5.0

δD(‰)

-20

Kushiro River

-40
-60
-80

-100

Fig3.Relationship betweenδ18O and δD in Hokkaido

Fig2.Sampling point around the Kushiro Moor
be contaminated and mixing of another place origin water byartificial influence.
Water was sampled with a 100ml polyethylene bottle withdouble seals and the value of pH, EC (electrical conductivity),
ORP (redox potential) and temperature values were measured
in the field using simple portable sensor.
3. ANALYTICAL METHODS
Dissolved major ions of sampled waters were measured byion chromatography, DX-ICS, AQ-1500 and ICS-1600(Dionex Ltd.). The measured cations are Li+, Na+, NH4+, K+,
Mg2+ and Ca2+. And the measured anions are F-, Cl-, Br-, NO3-,
PO43-, and SO42-.
The sampled waters were passed through the 0.45μm pore
diameter membrane filter in order to remove impurities suchas organic matter. Carbon dioxide (CO2), carbonic acid
(H2CO3), bicarbonate (HCO3-) and carbonate ion (CO32-),
such as all carbonate material in water samples was measured
by an acidic titration using 0.02N sulfuric acid.
The oxygen and hydrogen stable isotope ratios of water
samples were measured by the mass spectrometric systemusing the CO2-H2O and H2-H2O equilibration technique withplatinum as a catalyst. The mass spectrometric systems werecomposed of Finnigan Mat Delta Plus and WECs +GEO20-20 mass spectrometer. Oxygen isotope ratios aredefined as the 18O / 16O and hydrogen isotope ratios aredefined as D / H (2H / 1H). δ18O and δD values are defined as
the following equation (1).

  ( R x / Rst  1)  1000

δD=8×δ18O+10
δ18O(‰)(1)

Rx：stable isotopic ratio of x，Rst：stable isotopic ratio of st

4. RESULTS AND DISCUSSION
4.1 Wide distribution of stable isotopic ratios
Relationship between oxygen and hydrogen isotope ratios forsampled water is shown in Fig.3. The oxygen isotope values
ranged from -14.0 ‰ to -5.3 ‰ and hydrogen isotope values
ranged from -92 ‰ to 18 ‰. The relationship between
oxygen and hydrogen isotopic ratios for precipitation,
meteoric line is defined by the equation (2)δ D  8 δ18 O  10

(2)

Most of precipitation is on the meteoric line but the sampled
water in the whole Hokkaido area was divided into twogroups shown in Fig.3. One was on the meteoric line and theother was separated from the meteoric line. The separationfrom the meteoric line was caused by origin of precipitation[6].
Fig.4 and Fig.5 show the distribution of the δ18O and δD in
the whole Hokkaido. δ18O and δD values were the nearlysame tendency in the north region and around the Kushiro
Moor. Low δ18O and δD values are distributed in thenortheast region and the boundary between the northeast andthe central region. In the central region, the Tokachi
Mountains and high altitude areas show low values.
Generally, δ18O and δD values decrease with increasing
altitude and then it is called altitude effect. And δ18O and δDvalues decrease with increasing distance from the coast andthen it is called inland effect.
4.2 Estimation of the spring waters origin in the Kushiro
Moor
Relationship between oxygen and hydrogen isotopic ratios
for the Kushiro area is shown in Fig.6. The oxygen isotope
values ranged from -12.0 ‰ to -7.0 ‰ and hydrogen isotope
values ranged from -79 ‰ to 43 ‰. Fig.6 shows the linear474

relationship between δ18O and δD. Fig.7 shows the
distribution of the δ18O values around the Kushiro Moor andthe detail distribution of spring points. The oxygen isotope
values ranged from -12.0 ‰ to -7.0 ‰. The most of water
values ranged from -10.0‰ to -9.0‰ around the Kushiro
Moor. Increasing distance from the coast, δ18O values
decrease about 1.0‰. The δ18O values for the Kushiro Moor
■
○
▲
□
●
△

were thought to decrease about 1.0‰ with increasing distancefrom the coast by the inland effect.
Relationship between δ18O values and altitude is shown in
Fig.8. The δ18O and δD values decreased with increasing
altitude. The isotope altitude effect of the Kushiro Moor is-0.23‰/100m for δ18O and -1.37‰/100m for δD by theprevious study [7]. The altitude effect result of this study is-15.0‰ to -13.0‰-13.0‰ to -11.0‰-11.0‰ to -9.0‰-9.0‰ to -7.0‰-7.0‰ to -5.0‰-5.0‰ toFig4. Distribution of oxygen stable isotopic ratios

■
○
▲
□
●
△
▽
▼
×
＋
◆

-200‰ to -180‰-180‰ to -160‰-160‰ to -140‰-140‰ to -120‰-120‰ to -100‰-100‰ to -80‰
-80‰ to -60‰
-60‰ to -40‰-40‰ to -20‰-20‰ to 0‰0‰ toFig5. Distribution of hydrogen stable isotopic ratios475

-0.26‰/100m for δ18O and -2.02‰/100m for δD.
The difference between both results was thought to be causedby difference of sampling season and sampling points.
The right side figure shown in Fig.7 shows the spring waters
around the Chiruwatunai River. The spring waters areclassified into three groups. The first group consists ofS-01(-10.3‰) with relatively low δ18O values and is locatedaround the top of the Kirakotan Cape. Second group consistsof S-16(-8.5‰) and S-35(-8.1‰). The second springs arelocated in the upper Chiruwatunai River, and show highvalues compared with stream water around there. Third group
-13.0
-40

-12.0

-11.0

-10.0

-9.0

-8.0

consists of S-34(-6.9‰) and S-36(-7.1‰). The third springs
are located in same area of second group but show very highvalues. The δ18O values of the stream water around there areabout 9.3‰.
The first springs group with oxygen isotopic ratio, about-10.0 ‰ was thought to be derived from the forest area at thenorth of 10 to 20km from the Kushiro-Moor because theisotopic ratios of the stream water at the A area shown in theFig.7 were in agreement with those in the spring water.
Then, Fig.8 shows Hexa diagram representing major ionsconcentration to compare the spring waters and river water.
700

-7.0

600
500
400

Altitude(m)

δD(‰)

-50

-60

-70

300
200
100
0

δD=8×δ18O+10

-80

-13.0

-12.0

-11.0

Fig6.Relationship betweenδ18O andδD around
The Kushiro Moor

■ -12.0‰ to -11.0‰○ -11.0‰ to -10.0‰▲ -10.0‰ to -9.0‰□ -9.0‰ to -8.0‰● -8.0‰ to -7.0‰△ -7.0‰ to

-10.0

-9.0

-8.0

-7.0

δ18 O(‰)

δ18O(‰)

Fig8. Relationship betweenδ18O and Altitude

S-01(-10.28)
Kussyaro Lake
Akan Lake

Kirakotan CapeA

S-16(-8.40)Kushiro River

Tiruwatsunai River
S-34(-6.92)

S-35(-8.07)
S-36(-7.19)B
SpringKushiro MoorC
Akkeshi Lake

Fig7. Distribution of oxygen stable isotopic ratios around the Kushiro Moor
and Spring point around the Chiruwatunai River
476

Kushiro River

Sampling
name(δ18O:‰)

As both the A area and the first spring group (S-01) havesame character with low values of Mg2+ and high values of
HCO3- and Na+ + K+, the first spring group is also thought tobe derived from the A area.
Although the character of major ions for the A area and thefirst spring water is same, total concentration of the springwater is higher than those of the A area because generally
recharge water is added by ions from sediments or rocks
during migration.
The stream waters with oxygen isotopic ratio of the second
spring group were found in two areas, the B area and C area
near the Toro Lake located on the east side of 5km or 6km

away from the Kushiro Moor shown in Fig.7.
The second spring group (S-16 and S-35) was divided by into
two groups by the main ions. S-16 is characterized by highvalues of HCO3- and Ca2+ and is similar to those of all stream
water in the B area and some river water in the C area. On theother hand, S-35 is characterized by high values of HCO3- andNa+ + K+ and is not similar to stream waters in both the B and
C area.
The third group with very high values about -7.0‰ wascharacterized by high values of HCO3- and Na+ + K+.
Although the isotopic ratio for the third spring group is high,its Cl- concentration is low and then sea water mixing is not(meq/L)
2

1

0

1

SOSO44(meq/L)

2

2

1

Mg
Mg

SO4

Ca
Ca

HCO3
HCO
3

1

(meq/L)2

2(meq/L)

2

Mg

Ca

Cl+NO3 Cl

Na+K

Na+K

S-16(-8.40)

S-35(-8.07)

0

1

HCO3

Cl+NO3 Cl

S-01(-10.28)

0

SO4

Ca

Na+K
Na+K

1

Mg

HCO3

Cl+NO3 Cl
Cl+NO
3 Cl

1

0

(meq/L)

1

1

0

1

(meq/L)
1
SO4

Mg

SO4

0

Ca

HCO3

Na+K

Cl+NO3 Cl

K41(-10.10)

Mg

SO4

Mg

1

0

Mg

SO4

HCO3

Na+K

Ca

Cl+NO3

KE21(-8.31)(meq/L)

1

SO4

Ca

Cl+NO3 Cl

K43(-10.30)

0

1

Na+K

(meq/L)
1

0

Ca

HCO3
Cl+NO3 Cl

1

Mg

SO4
HCO3

(meq/L)

1

Na+K

Cl

KE18(-8.09)

1

B

Mg

(meq/L)
1
HCO3

Ca

HCO3

0

Na+K

Cl+NO3 Cl

K54(-10.10)

HCO3

K45(-10.50)

0

SO4

HCO3

1

2

Cl+NO3 Cl

S-34(-6.92)

0

SO4

Ca

HCO3

Na+K

0

SO4

Na+K

1

1

2

3

Mg

Ca

Na+K

S-36(-7.19)
Fig9. Hexa diagram : Sampling name(δ18O:‰)

477

Ca

Cl+NO3 Cl

Na+K

KE24(-8.18)
C

Mg

Cl+NO3 Cl

HCO3KE23(-8.12)(meq/L)
1

Mg

1

Mg

Ca

Cl+NO3 Cl(meq/L)
1

2

Na+K

A

2

3

Ca

SO4
Cl+NO3 Cl(meq/L)1

thought to increase isotopic ratio for the third spring.
Although the river water with -7.0 ‰ outside of the Kushiro
Moor was not found, Stream water with high isotopic ratio
may exist in the southeast and southwest in the Kushiro Moor
because there is an inadequate area for sampling and theprevious study showed high isotopic ratio in there.

[5]
[6]

[7]

5. CONCLUSIONSThe precipitations in the whole Hokkaido area were divided
into two groups by oxygen isotopic ratio. One was on themeteoric line and the other was separated from the meteoricline. The separation from the meteoric line was caused by thecondition of precipitation origin.
Oxygen and hydrogen stable isotopic ratios in the northregion and around the Kushiro Moor were high.
Low δ18O and δD values are distributed in the northeast
region and the boundary between the northeast and the central
region. In the central region, high altitude areas show low
values. The δ18O values decrease with altitude and distancefrom coast and then altitude effect and inland effect of isotope
can be observed.
Low δ18O values and high δ18O values spring waters were
found in the Kushiro Moor and there were classified into
three groups by oxygen isotopic ratio.
The first spring group with oxygen isotopic ratio, about-10.0 ‰ was thought to be derived from the forest area at thenorth of 10 to 20km from the Kushiro-Moor. Because theisotopic ratios of the stream water at this area were inagreements with the spring water. And both of them havesame character with low values of Mg2+ and high values of
HCO3- and Na+ + K+.
The stream waters of the second spring group, about 8.0‰
were found in two areas, near the Toro Lake located at theeast of 5 km or 6 km away from the Kushiro Moor.
Major ions of the one spring in the second group are similar to
those of stream water in those areas. On the other hand, major
ions of the other spring in the second group are not similar to
stream waters in those areas.
The third group with very high values about -7.0‰ wascharacterized by high values of HCO3- and Na+ + K+.
Although the river water with -7.0 ‰ outside of the Kushiro
Moor was not found, stream water with high isotopic ratio
may exist in the southeast and southwest in the Kushiro Moor
with inadequate for sampling.
6. REFERENCES
[1]
[2]
[3]

[4]

Kushiro Nature Restoration Committee, “Comprehensive Plan Naturalregeneration Kushiro,” 2005
Kenzo HIROKI, “Integrated Water Resources Management for better
groundwater management,” Journal of Japanese Association of
Hydrological Sciences, Vol.40, No.3, 2010, pp.85-93
Masahide Ishizuka,Yumi Sone, Hiroyuki Ii and Tatemasa Hirata,“Effect of enriched early rainwater on mesoscale isotopic distribution
in surface water on the Kii Peninsula, Japan,” Water Resources
Research, Vol.42, W12410, doi:10.1029/2004WR003810, 2006
Kyohei Yokota, Hiroyuki Ii and Masanobu Taniguchi,”Effects ofaltitude and distance from castline of oxygen stable isotope for
precipitation in the Yakushima Island,” Environmental Engineering
Research,Vol.47, 2010pp.667-677

478

Editorial Board of the Hokkaido region, “Geology of Japan 1 Hokkaido
Region,” Kyoritsu publication, 1990
CHITOSHI MIZOTA and MlNORU KUSAKABE, “Spatial
distribution of 6D-6180 values of surface and shallow ground watersfrom Japan, south Korea and east China,” Geochemical Journal,
Vol,.28, 1994, pp.387- 410
National Institute for Rural Engineering, “Estimation of groundwater
rechargearea
asmeasuredbyisotopic,”
http://www.nkk.affrc.go.jp/library/newpanel/pdf/nouchimizu/panel%2
010.pdf

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

A Debris Flow and Its Risk Analysis Related to the 2008 Wenchuan Earthquake
Yange Li1, Guangqi Chen2, Chuan Tang3, Lu Zheng1
1

Department of Civil and Structural Engineering, Kyushu University, Fukuoka, Japan
Department of Civil and Structural Engineering, Kyushu University, Fukuoka, Japan & Visiting Professor of
Institute of Earthquake Science, Beijing, 100036, China
3
State Key Laboratory of Geo-Hazard Prevention, Chengdu University of Technology, Chengdu, China

2

ABSTRACT: On August 14, 2010, an intense and prolonged rainfall
triggered debris flow in Hongchun gully, Yingxiu town, theepicenter of the Wenchuan earthquake. The event transported hugematerial sediment to the watercourse of the Minjiang River andformed a natural dam, resulting in flash flood in the newreconstruction Yingxiu town. In order to predict the run out distanceand extent of the hazard area of such kind of debris flows, wedevelop an approach from the earthquake-induced landslides
recognition to the numerical simulation of the debris flow in thisstudy. At first, we determined the initial material sediment by aerial
photography identification and field investigation. And then, a
two-dimensional numerical model was used to simulate the transport
process of the debris flow. The raster grid network of digital
elevation model generated from GIS was used for the finite
difference mesh. The results from the numerical simulation showed
a good agreement with the actual movement and deposition
distribution of Hongchun gully debris flow.

Keywords: debris flow; aerial photography identification;
two-dimensional numerical model; GIS
1. INTRODUCTION
The Wenchuan earthquake (Ms=8.0), with a focal depth of
12-15 km, throughout an area of about 50, 000km2, wastriggered by the reactivation of the Longmenshan fault(LMSF) zone at eastern margin of the Tibetan Plateau,adjacent to the Sichuan Basin on May 12th, 2008. The
earthquake led to more than 69,225 fatalities, 374,640
persons injured, about 17,393 persons missing, 5,362,500
collapsed homes and 21,426,600 homes that were badly
damaged by April 25th, 2009 according to official statics (Cuiet al., 2009, Tang et al., 2011a).
The latest data from aerial photographs and satellite images
shows over 56,000 landslides occurred after the earthquake(Dai et al., 2011). The majority of landslides wereconcentrated in the hanging wall part of the
Yingxiu-Beichuan fault and Pengguan fault, northwest part ofearthquake zone. Consequently, a great deal of loose
sediment was induced which in turn promoted heavy debris
flows during subsequent typhoons and heavy rains. The mostoutstanding example was a severe debris flow hazard
occurred in Hongchun gully, Yingxiu town, epicenter of the
Wenchuan earthquake on August 14, 2010. The event
transported huge sediment material to the water course of the
Minjiang River and produced a natural dam, which thenchanged the course of the river and resulted in flash flood in
the new constructive Yingxiu town. The flood claimed 13

lives, with a further 59 listed as missing. About 8,000 local
residents were rapidly evacuated from their homes.
More recently, a giant debris flow burst on August 8, 2010 inZhouqu City, which is also in the earthquake-affected area,that claimed 1,434 lives with a further 331 listed as missing.
This catastrophic debris flow was following an intense local
rainstorm and led to the destruction of more than 5,500
houses along its flow path. All the events and previous
researches (Cannon et al. 2001; Dong et al.2009; Tang et al.,2011b) showed strong earthquakes play a major role incontributing to the accumulation of sediment supply on
hillslope and in channels. Bovis and Jakob (1999) and Jakobet al. (2005) highlighted the importance of sediment supply
on the frequency and magnitude of debris flows.
In order to predict the run out distance and extent of thehazard area of such kind of debris flows, we develop an
approach from the investigation of earthquake-induced
landslide to the numerical simulation of the debris flow in thisstudy. At first, we determined the initial material sediment by
aerial photography identification and field investigation. Fourlarge scale landslides were identified by using theobject-based remote sensing technique and they were thoughtas the main loose material of the debris flow, their volume
was estimated by the scale of the landslides. And then, weused a two-dimensional numerical model to simulate thetransport process of the debris flow. In the model, the debris
and water mixture was assumed to be a uniform continuous,
incompressible and unsteady Newtonian turbulence fluid.
The raster grid network of digital elevation model generated
from GIS was used for the finite difference mesh.
2. STUDY AREA AND DATA SOURCEThe Hongchun gully is situated on left bank of MinjiangRiver and in Northeast region of the Wenchuan County,
Sichuan Province, China (Fig. 1). The study area is situated in
the transitional mountainous belt between the Sichuan Basinand the Western Sichuan Plateau and the gully-mouth islocated at 31°04’01”N and 103°29’33”E. The geologicalmain structure and the strike of the rock strata in this area
show a NE-SW orientation. The Yingxiu-Beichuan fault justruns through the Hongchun gully.
The study area is situated in the typical humid subtropical,
monsoon climate zone. The annual average precipitation overa period of 30 years is 1,253 mm, with a highest recorded
annual precipitation of 1,688 mm in 1964. On average, 70%of the annual precipitation is largely concentrated in the
period from June to September. The maximum recorded
rainfall intensity was 269.8mm/day in 1964.

479

Fig. 3 Aerial photograph of Hongchun gully

Fig. 1 Hongchun gully of Yingxiu town
2

The Hongchun gully has a catchment area of 5.35 km and achannel length of 3.55 km. The upslope elevation of the gully
is more than 1,700m asl and the gully mouth is at 880m asl,
The highest part of the debris flow source area is at 2,168 m
asl. In the source area, mainly granitic rocks are exposed,
which are deeply fractured and highly weathered, andmassive coseismic landslides were developed on the slope.
On August 14, 2010 at 3:00 A.M., triggered by an intense
rainfall, numerous debris flows occurred near Yingxiu. The
Hongchun gully debris flow was the most catastrophic one. It
deposited mud, debris, and rocks onto the G213 national roadand the Dujiangyan-Wenchuan highway. The alluvial fan ofdebris flow was located on the region of under-construction
transition settlement building of Yingxiu town. It travelledacross Minjiang River and caused a natural debris dam of 10
m height, 100 m in length across the river, and 150 m in widthalong the river. The dam caused a change of the river courseand flooded the newly reconstructed Yingxiu Town (Fig. 2).

Fig. 2 The blocked Minjiang River and new Yingxiu town inflood by Hongchun gully debris flow
The basic data utilized in the study included a 20m × 20m grid
digital elevation model (DEM) and an aerial photograph witha spatial resolution of 0.3m (Fig. 3), which was taken on May
20, 2008 by the Ministry of Land and Resources (MLR) after
the Wenchuan earthquake occurred.

3. CAUSE ANALYSIS OF THE HONGCHUN GULLYDEBRIS FLOW
Debris flows are triggered by a combination of three essential
factors: sufficient available loose material, surface runoff,and steepness of the drainage channels on the slopes(Takahashi 1981). Therefore based on field investigation, thecause of debris flow in Hongchun gully can be explained as
the following 3 factors:
3.1 Triggering rainfall
Debris flows usually occur at a combination of highaccumulated rainfall and very high rainfall intensity (Crosta
1998). On 12 August a total of accumulated rainfall of 162.1
mm in 33 h was recorded from 17:00 P.M. until 02:00 A.M.
on 14 August. The rainfall intensity that induced the debris
flows occurred between 02:00 and 03:00 A.M.; 16.4 mm/h
rainfall intensity was recorded during that time period.
However, the peak hourly rainfall was 19.9 mm recorded
between 17:00 and 24:00 hours two days before the landslide.
Previous studies in the Longmenshan area (Tan and Hen
1992) have reported rainfall that initiates debris flows to haveintensities greater than 30-50 mm/h with a total rainfall of atleast 80-100 mm. Lan et al. (2003) examined rainfall amount
110 mm/day as the rainfall threshold for the occurrence of
past landslide events in the Xiaojiang watershed, south
western China. Comparing the triggering rainfall parameterwith the rainfall condition in Hongchun gully, it indicates that
although the antecedent precipitation of this event is higher
than record, the trigger rainfall intensity is about 50% reduced.
After the Wenchuan earthquake, debris flow susceptibility is
increased by a large precipitation event.
3.2 Sediment supply conditions
The comparison between pre-earthquake and post-earthquake
cases indicates that the strong earthquake has modified theconditions and debris flow is more easily induced after
earthquake by a smaller rainfall (Lin et al. 2003; Tang et al.
2009). It can be observed by remote sensing analysis of theaerial photographs taken on May 18, 2008, after the 2008
Wenchuan Earthquake that an abundance of landslide debris
was presented on the slopes. Therefore, the slopes are very
susceptible to debris flows under heavy rainfall conditions.

480

3.3 Topography
The average vertical of the Hongchun gully is 358‰, and the
upstream is 538‰. Slope steepness is a fundamentally
important factor influencing the occurrence of debris flows
because most debris flows initiate as shallow landslides or as
eroded rill or gullies on large landslide deposits. Since theoriginal area and pass channels of the Hongchun gully arestraight, the confluence of rainfall water became quick. The
highest part of the debris flow source area is at 2,168 m asl.
and the gully mouth is at 880m asl. The large elevationdifference gave the rainfall water large kinetic energy
transformed from potential energy to initiate the loose
sediment material to failure
4. LANDSLIDE RECOGNITION
Landslide inventory is the most important procedure forrelated disaster prevention. Landslide mapping by field
investigations is a challenging task in vast and inaccessible
mountainous terrain. Visual interpretation of remote sensing
data is time and labour consuming. So far there have only
been a few attempts at automating the mapping of landslides
by pixel-based methods, which likely fail as DN values alone
and do not characterise geomorphic processes such as
landslide (McDermid and Franklin, 1994). Recently Barlow
et al. (2006) achieved good detection accuracy by treatinglandslide as an object and only considering the knowledge
that landslides are quite large. Moine et al. (2009) used highresolution earth observation data without DEM, could
recognize small landslides, eventually ruling out thepossibilities of classifying landslide types. Tapas R. Martha etal. (2010) used the algorithm developed in Definiens
Developer software based on high resolution satellite data
and a DEM, achieved 76.4% landslide recognition accuracy
in five different landslide classed in a terrain featuring
spectrally identical land use/cover units. In our research, amulti-step approach was presented as following to recognize
and classify landslides accurately.
4.1 Image segmentation
An important step before characterising diagnostic attributesof features of interest, such as landslides, is the creation of
objects/segments that alone or in a group demarcate theboundary of the given feature. This is done using image
segmentation, which is a process of dividing the image into
objects or regions based on the homogeneity of the pixel
values. These segments ideally correspond to real-world
objects. Image segmentation can be done in different ways,
using techniques such as density slicing, split and merge(Kerle and de Leeuw, 2009).
In this research, we employ an edge-based segmentation
algorithm which is very fast and only requires one input
parameter, scale level. It is the regularization parameter, if it
is small, then a lot of boundaries are allowed and a “fine”
segmentation results. As it is increases, coarser and coarser
segmentations result. By suppressing weak edges to different
levels, the algorithm can yield multi-scale segmentation
results from finer to coarser segmentation.
Landslides pose a particular challenge to image segmentation,
as land cover variability (e.g. partial vegetation), andillumination variations as a function of terrain characteristics,
often result in spectrally diverse features. It is not practical toattempt outlining landslides as single segments for the firsttime, and some post-segmentation merging processing isneeded also due to the typical size variability of landslides in
an image (Martha et al., 2010).
We partitioned the image into segments by groupingneighbouring pixels with similar feature values. In order toensure the features of interest are not grouped into segments
represented by other features and to ensure that a feature ofinterest is not divided into too many small segments(over-segmentation), we firstly choose a small scale level 30,
to avoid the wider boundaries between segments which somefeatures of interest will lose.
Then segmentation merging is processed to resolve the oversegmentation problem. It used the Full Lambda-Schedulealgorithm which is created by Robinson, Rdding and Crisp(2002). The algorithm iteratively merges adjacent segments
based on a combination of spectral and spatial information.
Merging proceeds if the algorithm finds a pair of adjacentregions, i and j, such that the merging cost t i , j is less than adefined threshold lambda value:

Oi  O jti , j 

Oi  O j ui  u j

2(1)

legnth ( (Oi , O j ))Where: Oi is region i of the image; Oi is the area of region i;

u i is the average value in region i; u j is the average value inregion j; u i  u j is the Euclidean distance between the
spectral values of regions i and j; legnth( (Oi , O j )) is thelength of the common boundary of Oi and O j . In thisresearch, we find the merge scale 93 can best represent thelandslide body. Fig. 4 shows the segmentation result.

Fig. 4 Image segmentation result
After segmentation, the spatial, spectral, texture and colour
attributes were calculated for each object.

481

4.2 Image classification
Bare rock or debris exposed after the earthquake, giving abright appearance in the image, although at times mixed withremaining or dislodged vegetation. The Normalized
Difference Vegetation Index (NDVI), a simple graphicalindicator that can be used to analyse remote sensing
measurements, typically but not necessarily from a space
platform, and assess whether the target being observed
contains live green vegetation or not, is sensitive to low levelsof vegetation cover, has been successfully used by previous
works (Barlow et al., 2006; Schneevoigt et al., 2008) todiscriminate landslides from vegetated features. Therefore,
we used NDVI as a first criterion to identify landslide
candidates, and separate them from other areas such as forest
land, orchards and crop land. An NDVI of 0.15, a value close
to the statistical mode of the image NDVI, was found to be
useful for discriminating landslide candidates fromvegetation cover. Since NDVI is used as a cut-off criterion,
objects with similar or lower than the NDVI threshold, suchas rock outcrops, roads, water bodies and river beds, are likely
to be classified as landslides candidate (Fig. 5).

in n-dimensional space of the target to the elements in the
training data, where n is defined by the number of objects
attributed used during classification. This method is generally
more robust than a traditional nearest-neighbour classifier,since the K nearest distances are used as a majority vote todetermine which class the target belongs to. The K Nearest
Neighbour method is much less sensitive to outliers and noisein the dataset and generally produces a more accurateclassification result comparing with traditional nearest
neighbour methods. In this case, we define the K Parameter as
3 that it means there are 3 neighbours will be consideredduring classification. Fig. 6 gives the classification result.

Fig. 6 Image classification of Hongchun gully

Fig. 5 Landslides candidate recognized by NDVI calculation
In order to classify the landslide candidates, the training
samples need to be assigned. Different landslide type showsdifferent spectral characteristic. For example, comparing withrocky landslides, debris flow shows relatively darker
appearance. So we divide the landslides into three categories
when selecting the training samples. Meanwhile, considering
the landslide and some resident have similar spectral features,in order not to miss the landslide area, we take theconservative thinking which is not defining the resident areain the image at first. Then all the landslides will be recognized
as well as some mis-classified resident area included. Finallythe three kinds of landslides were merged within ArcGIS.
The more features and training samples are selected, thebetter the results are from supervised classification. However,selecting an overwhelming number of training samples willcause poor performance during classification and previewingclassification results. In our case, 30 rock slides, 38 debris
flows, 30 debris slides, 10 river objects were selected in total
as the training samples for supervised classification.
The K Nearest Neighbour algorithm was used for theclassification. The algorithm considers the Euclidean distance

4.3 Post classification
As shown in Fig.6, we can see almost all of the landslides
were identified, although some resident area was also
included. In order to improve the classification results, thenon-landslide objects are eliminated by the assumption thatlandslide will not occur for the slope gradient less than 9° and
it is processed in ArcGIS. A total of 73 landslides weresemi-automatically detected in the entire area, with a surface
area of 1.0 km2 in the Hongchun catchment area.

Fig. 7 Landslide recognition of Hongchun gully
Meanwhile, we can see some objects with blue colour weremis-classified as river, they were rectified manually (Fig. 7).

482

5. TWO-DIMENSIONAL NUMERICAL MODEL OF
DEBRIS FLOW
Based on the conservation of mass and the momentum of theflow, many researchers have proposed mathematical modelsof the debris flow. Some of them are 2D models (Takahashi etal., 1992; Chen and Lee, 2000; Ghilardi et al., 2001;
Denlinger and Iverson, 2001; Wang et al., 2006). In this study,
the debris and water mixture is assumed to be a uniform
continuous, incompressible, unsteady Newtonian fluid. The
numerical solution using a finite-difference formulation
based on the DEM grid was detailed depicted in the previous
literature (Wang et al., 2006).
The proposed approach has been used to simulate the
Hongchun gully debris flow. Furthermore, the grid networks
in GIS are used as the grids of finite-difference method.
When making DEM analysis in GIS, each cell has eight
possible flow directions (Fig. 8). The flow direction of a cellis expressed in degrees: left = 0, up = 90, right = 180, down =270; and the diagonals: 45, 135, 225 and 315. Within a cell,
overland flow is routed along one flow direction. The flow
direction is the maximum downslope direction, which isdetermined from the raster-based DEM.

Fig. 9 Loose material identified for the debris flow
The movements of the debris flow are illustrated in Fig.
10(a)-(d) for different time. It can be seen that the river isblocked in Fig. 10(d). The distribution of the maximum depthof the whole flow is shown in Fig. 11.

Fig. 8 Possible flow direction in a grid cell
Since only a 20m DEM map is available for generating grids,
the area is divided into 200×170 grids by using ArcGIS9.3.
The rheological parameters are assumed as constant (Tang etal., 2011b) and listed in Table. 1.
ρd2050 kg/m3

Table. 1 Rheological parameters
α
β
μ
1.25
1.0
0.11

Fig.10 The movement of the debris flow

tanφ
0.6

where, ρd is the density; α and β are the momentum correct
factors; μ is the kinematic viscosity; and tanφ is the friction.
The photo interpretation can only provide information on thelandslide surface area, so a relational model between thelandslide area and its average thickness is required todetermine the volume of the landslide debris. Combining with
the identification result and field investigation, we finallyselected four landslides: H1, H2, H3, H4 as the main loose
source material of the debris flow (Fig.9). Information
concerning coseismic landslide thickness was derived fromthe data in the report: “The investigation report on debris flow
hazard in the Hongchun gully of the Yingxiu town after the
Wenchuan earthquake”, published by the Guanghan Instituteof Geological Engineering Investigations of Sichuan
Province. The area for the above 4 landslides are separately
7,688 m2, 5,137 m2, 2,002 m2, 4,567 m2 , and there average
depth are separately defined as 4m, 3m, 1m and 1m.

483

Fig.11 The distribution of the maximum depthof Hongchun gully debris flow[4]6. CONCLUSION
A strong earthquake not only cause directly damage on
constructs but also can result in a series of natural disasters
such as landslide, debris flow and flooding. These secondary
disasters occur as chain disasters after a strong earthquake.
For mitigation of such kind of risk, apart from engineering
measures, we have proposed an approach from the earthquake
induced landslides investigation to the numerical simulationof the debris flow.
In the first stage, a multi-step segmentation approach has
been used to recognize and classify landslides accurately.
Firstly, we divided the image into objects based on
homogeneity of pixel values through edge-based
segmentation algorithm. Then to ensure the features ofinterest are not over-segmentation, the merge segmentation
based on the Full Lambda-Schedule algorithm was used to
improve the delineation of feature boundaries. After
segmentation, spatial, spectral, texture and colour attributes
were calculated for each object. Next NDVI index was used
to separate the landslide candidates from vegetation. Finally,
supervised classification using the K Nearest Neighboralgorithm was applied to the entire image and themis-classified objects were rectified by expert-driven
knowledge. Another achievement of this study was detectionof complex failure mechanism for different kinds oflandslides. Considering the spectral characteristic for thelandslides in different sites shows different appearance in theimage, such as debris flow shows relatively darker brightness
than rocky landslide, we separately chose the landslide
samples in three categories according to the visual spectral
value, and last they were merged together after theclassification. Results showed almost all of the landslides
were identified as well as some resident area included.
Based on the image interpretation, a depth-averaged 2D
numerical model incorporating the grid in ArcGIS used as thefinite difference mesh was employed to simulate the debris
flow in Hongchun gully. The simulation can show us a realpropagation process of the debris flow, the potential
inundation area can be derived and the maximum depth in theflow gully was calculated.
7. ACKNOWLEDGMENT
This study has received financial support from Grants-in-Aid
for Scientific Research (Scientific Research (B), 22310113, G.
Chen) from Japan Society for the Promotion of Science andthe Research Fund of the State Key Laboratory of
Geo-Hazard Prevention of China (SKLGP2009Z004). The
financial supports are gratefully acknowledged.
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Remaitre A & Bogaard, T (Eds.), Landslide Processes: From
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First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Impact Analysis of Water Price Reform of Zhangye, China
Atsushi Koike, Dept. of Civil Eng., Kobe University, Japan,
Zhongmin Xu and Kang Wang, CAREERI, Chinese Academy of Sciences, China
Bunei Itoga, Regional Futures Research Center, JapanABSTRACT: The Hei River Basin is located in northwest region ofChina. This region belongs to arid zone, and water resources are oneof the main limiting factors of harmonizing the development of
ecology, economy and society. Due to the economic growth and the
population increase, water consumption has grown rapidly since
1970s in the middle reaches especially in the Zhangye district. The
rapid increase of water demand in the middle reaches has degraded
the ecosystems of the whole watershed. Water pricing is very
important instrumentally in balancing water supply and demand.
Appropriate water price is a very effective countermeasure inchanging the social behavior towards water resource conservation,
promoting economic efficiency and investment in more efficientequipments. The aims of this study are to analyze the comprehensive
impacts of Water Price Reform of the Zhangye district by using an
operational Computable General Equilibrium (CGE) model basedon building a social accounting matrix (SAM).

Keywords: Water Demand Management, Water Price Reform,
Computable General Equilibrium model, Zhangye, China
1. INTRODUCTION
The Hei River Basin is one of the major grain producing
regions in China. The water shortage is mainly caused by the
drastic population growth and the development of theirrigation area in the middle basin over the past decades.
Water saving measures have to be taken to improve theirrigation system efficiency both on-farm and in the main
canal system including water-saving measures, efficiency
improvement and reuse of water. On the other hand,institutional management, water pricing and agricultural
sector adjustment will play an important role in balancing
demand and supply [15].
Water management includes water supply management andwater demand management (WDM). WDM refers to theactivities that aim to reduce water demand, improve water use
efficiency and avoid the deterioration of water resources.
Demand management offers sustainable water management
solutions in the face of increasing water scarcity and growing
conflicts over water use [6], [12]. For the Hei River Basin, theWDM measures consist of both technical and non-technicalmeasures. Since irrigated agriculture uses 80% of theavailable fresh water, and the efficiency of this water use isvery low, large quantities of water are wasted. Considerable
economies can be realized if agricultural water is used in amore efficient way. Furthermore, water pricing andinstitutional reform can help to control water demand andimplement water saving [15].
Appropriate pricing of water (i.e., implementing an
increasing block rate pricing structure) has proven to be avery effective measure in changing the public behavior
towards water conservation and promoting economic
efficiency and investment in new equipment [10].
However, the reform of water price will take impacts inregional economic system. The data on direct water
consumption reveal that the amount of water consumed
directly by the primary industry (agriculture, forestry,
livestock, and fishery) is much greater than that consumed by
the industrial and service sectors, with agricultural
consumption exceeding 2 billion m3, and the latter only
consuming a small fraction of this amount (approximately 55
million m3). This finding confirms the well-known fact that
agriculture is the main consumer of water resources in
Zhangye district, and is responsible for 94% of the total water
consumption in the region. In comparison, the volume ofwater consumed directly by the industrial and service sectors
is nearly negligible. However, when indirect water isconsidered, it becomes obvious that water consumption by
the industrial and service sectors increased greatly. This isoften unnoticed in analysis that focuses exclusively on thelower values for direct water consumption by these sectors.
This means that although these sectors use only a small
amount of water directly in production, in order to produce
the inputs (generated by other sectors) that they incorporate
into their production processes, a high consumption of wateris necessary. Thus, it appears that the industrial and service
sectors also consume large amounts of water indirectly. In
this sense, indirect consumption seems to make up asignificant part of the water consumption in the study area [17].
Therefore, water pricing will impact not only the agricultural
sector but also other sectors.
Water resources are the fundamental components which drive
the evolution of ecological-economic system. Therefore, weshould evaluate the impacts of water price reform before it
was implemented. The aims of this study are to analyze thecomprehensive impacts of Water Price Reform of theZhangye district by using an operational Computable General
Equilibrium (CGE) model based on building a Social
Accounting Matrix (SAM).
In this study, we build the SAM for water price reform and the
CGE model based on the SAM. And, three water price reform
scenarios are considered. We present the results fromsimulation experiments that we performed with the model toanalyze the effects of each water price reform scenario.
2. STUDY SITE

The Hei River Basin spans Qinghai, Gansu and Nei Mongol,
and is located in the arid zone of northwestern China. This is
the second largest inland river basin in China. It covers an
485

area of approximately 130,000 km2. Its upper reaches sourcefrom the boundary district of Gansu and Qinghai, and its
lower reaches end to the desert in the western part of Inner
Mongolia. Administratively, the basin includes a county of
Qinghai Province located in the upstream region of the Hei
River Basin; a city and counties of Gansu Province, all ofwhich lie in the midstream region, namely Zhangye district,
Minle district, Shandan district, Linze district and Gaotai
district, respectively; and a county (within the Ejina Oasis
with the location in the downstream region of the basin) in the
Inner Mongolia [9] (Fig.1).The study site is Zhangye district,located in the middle reaches of the Hei River, is 42,000 km2
in size and has a population of 1.264 million, including a rural
population of 9.11 million and an urban population of 3.53
million. The climate of this region is arid, with annual
precipitation ranging from approximate 100 to 300 mm, andpotential annual evapotranspiration reaching 2,000 mm.
Although located in one of the driest zones in the world,
Zhangye district consists of many oasis ecosystems that are
mainly watered by the Hei River. Water use in this city
accounts for about 93% of all water use from the river, with
94% of this water used for agriculture. According to the
Zhangye Statistical Yearbook [4], the irrigated area in Zhangye
district was about 68,667 ha in the 1950s, but by 2002, it
expanded to approximate 266,000 ha, including 212,000 ha offarmland and 41,000 ha of forest and grassland. As a result of
irrigated farming, Zhangye district has become an importantcenter of Gansu Province for the production of commodity
grains.
Since the Chinese national economic reforms that began in
1978, new industrial sectors have arisen, such as mining(including coal production), production of building materials,
electric
power,
metallurgy,
machineryassembly,transportation, and services. In recent years, Zhangye district
has experienced considerable economic growth as a result of
these changes. The gross domestic product (GDP) was 837.3
million US$ in 2001, which was 8% greater than that in 2000.
In 2002, 2003, and 2004, the GDP increased to 916.8, 1013.1,
and 1206.7 million US$, respectively, representing annual
increases of 10%, 11%, and 12%, respectively, over thevalues in the previous year [8].
Expanding agriculture and rapid economic growth haveresulted in excessive use of the region’s water resources.
According to GPBWR (2003) [7], in 2002, the annual available
water resources were 2.05 billion m3, including 1.63 billion
m3 surface water and 0.42 billion m3 groundwater while theactual annual water utilizations were 2.42 billion m3, of which
90% was consumed by the socio-economic systems, and of
this amount, 96% was used for agriculture. Ecological andenvironmental water demands are severely restricted for theexcessive water use in socio-economic systems. As a result,
the
city
seems
to
have
lockedintoanenvironmental-economic dilemma through increasing
dependency on the scarce water resources and further erosionof environmental quality [16].

Fig.1 Hei River Basin [15]3. CGE Model for Water Price Reform
3.1 Computable General Equilibrium model
A Computable General Equilibrium (CGE) model is a generalequilibrium model that implements the textbook descriptionof an economy. There are utility-maximizing consumers
whose decisions determine the demand for goods and supplyof labor. There are profit-maximizing producers whose
decisions determine the supply of goods and the demands forprimary factors (labor, capital, and land) and intermediate
inputs. There is international trade. There is a government
which collects taxes and tariffs; may set exchange rates; andprovides transfers, subsidies, and services. Finally, there aremarket-clearing conditions specifying supply-demandbalance, which will determine equilibrium prices. The model
is a "general equilibrium” because all domestic supplies,
demands, prices, and incomes are determined simultaneously
within the model. It is computable because the model solves
empirically for all endogenous variables in a highly
non-linear system of simultaneous equations.
Changes in policy alter demand through changes in prices.
The wide scope of the model makes it especially useful forevaluating projects that have broad effects, changing incomesin many sectors through intersectoral linkages. When there isa generating many ripples in the economy, a generalequilibrium framework are the appropriate tool of analysis [3].
The CGE model can consider the complex relationships ineconomic system, and it has been a popular tool in policy
analysis. Water issues are important also in the world.
Numerous state and regional economic impact studies ofwater management have been conducted. [Peter Berck, 1990]
employ CGE procedure to investigate the reallocation ofwater in the San Joaquin Valley [11]. and [Chang K. Seung,1997] investigate the economic impacts of transferringsurface water from irrigated agriculture to recreational use at
the Stillwater National Wildlife Refuge in Churchill Country,
Nevada[5]. [Alexander Smajgl, 2006] develop a conceptual
framework of water reform and generates an Applied General486

Equilibrium (AGE) model to investigate the impacts ofpotential water reform scenarios for an irrigation area withfeatures of the Lower Burdekin [1]. [Okuda and Hatano, 2005]
provides water-rights transaction model at China’s provincelevel applying general equilibrium theory, in which sets avirtual water-rights market for YRB [13]. We built the Social
Accounting Matrix (SAM) [2] of Zhangye district and used an
operational CGE model to analyze the impacts of water pricereform of Zhangye district. And this study analyzes theimpact on the Zhangye economy by water price reform
methods by using the CGE model.
The CGE model explains all of the payments recorded in the
SAM. The model therefore follows the SAM disaggregationof factors, activities, commodities and institutions. It is
written as a set of simultaneous equations, many of which arenonlinear.
3.2 Features of CGE Model for Water Price Reform
Fig.2 shows the Framework of the CGE model for water pricereform. Continues line shows flow of goods and factors, andit shows money flow in opposite sense. Dash line shows theflow of tax, subsidy, saving and transfer.
factormarketlabor, capitaltransferlabor, capital, water

water

capital, water authority
tax
investment
tax

government

householdinvestment
agent

savinggoods

goods
marketexporttransfer:agents:marketenterpriseinvestment

tax goods

goods

goods

It is supposed that government revenue is composed of value
added tax, household tax and direct tax of water authority.
Government expenditures are divided into transfer tohousehold, consumption and saving.
Each producer is assumed to maximize its profits, defined as
the difference between revenues earned and the cost. Profits
are maximized subject to a production technology; thestructure of which are shown in fig.3.
Value added (V) is composed of water (W), labor (L) andcapital (K) by the Cobb-Douglas function. Domestic
production (DP) is a combination of value added andintermediate (X), which are characterized as strict
complements according to Leontief function. In other words,
zero-value substitution elasticities are assumed forintermediate inputs and value added. It is assumed that input
coefficients of intermediate goods are fixed. Domestic
production is transformed to domestic goods (D) and export(E) by CET (constant elasticity of transformation) functionand export price is domestic consumer goods price. Domestic
consumer goods are obtained to compose of domestic goods
and import (M) by CES (constant elasticity of substitution)function and import price is fixed the price of rest of the world.
The import price is the price paid by domestic users forimported commodities. And, the export price is domestic
consumer goods price. In Zhangye district, most of importand export is transaction between Zhangye district and othernational region. In this model, we assumed that the importprice is fixed the price of rest of the world. In this model,
foreign saving and transfer to household from rest of theworld are fixed. The fact that all items except imports andexport are fixed mean that, in fact, the trade deficit also isfixed. Model detail is in appendix.

importrest ofthe world

domestic consumer goods
investment

CES:flow of goods and factors:flow of tax, subsidy, saving, investment, transfer

importdomestic goods

Fig.2 Framework of the CGE model for water price reform

ExportCET

In this CGE model, agents are presented by household, waterauthority, government, rest of the world, investment agent
and ten enterprises. And there are factor market and goods
market. Enterprises are “agriculture”, “forestry”, “livestock”,“fishery”, “agriculture, forestry and fisheries service(agriculture service) ”, “mining”,“manufacture”,“electricity”, “construction”, “others”.
The household income is composed of returns to labor andcapital, as well as transfers from government and the rest of
the world. And it is supposed that transfers to households
from government are proportionate to government income.
Household expend on saving, consumption of commodity andtax. The households consume the domestic consumer goods.
It is assumed that each household maximizes its utility
function subject to consumption expenditure constraint. The
utility functions are defined as Cobb-Douglas functions.
The water supply authority (water authority) holds water andcapital, and earns factor incomes. Water authority expend totax and saving

domestic production
Leontief
intermediate

value added

Cobb-Douglas
water

labor

capitalFig.3 Framework of Production Technology
3.3 Definition of Welfare Loss
The change in welfare is defined as the change in equivalentvariation (EV). The EV is an estimate of the hypotheticalvariation in the household income, which would haveproduced the same change in the utility of the representative
consumer, at fixed prices.
Amount of water saving is defined as the amount of waterusing without water price reform minus the amout of waterusing with water price reform.

487

Senario3
-3
0%

20%
40%
60%
80%
water saving rate with primary water price

Paw  P w

Pw

：primary water priceP：water price increase ratea：suffix of activity

OthersFishery

2%

0

0%

-5

-2%

-10

-4%

-15

-6%

-20

-8%
Scenario1 Quantity
Scenario1 Rate

：water price increase amountchange rate

Constant price for all activities
scenario3

quantity(million Yuan)

Paw  Paw  P

Livestock5

Constant additional price charge for all activities
scenario2

Forestry

Paw  1   Paw

Agriculture

Constant rate price charge for all activities

Construction

4.2 Impact on Domestic Production of Water Price
ReformPFig.6 shows the change of quantity and the change rate ofdomestic production. Fig.6 shows that water price reform of
scenario1 affects the domestic production of mostly
agriculture and manufacture. On the other hand, water pricereform of scenario 2 affects domestic production of mostly
agriculture. Even though, impact on agriculture of scenario1
is larger than scenario 2 although the water price charge for
agriculture of scenario1 are smaller than scenario 2.

Table 1. Scenario definition
scenario1

100%

Fig.5 Welfare loss with water saving rate

Electricity

Fig.4 Primary water price of each activity (Yuan/m3)

Senario20

Manufacture

Other

Construction

Electricity

Manifacture

MiningFishery

Agriculture
sarvice

Livestock

Forestry

Agriculture0

Senario1
3

Mining1
0.5

6Agriculture sarvice

1.5

9

Scenario2 Quantity
Scenario2 Rate

Fig.6 Change of quantity and change rate of domestic
production to save the water of 1 billion m3

4. RESULTS AND INTERPRETATION

Agriculture sarvice

2

when the water price is more than 0.7 (Yuan/m3) (continuousline in fig.5). Looking at the welfare loss with quantity ofwater savings, scenario 2 is the best water price reform
method except for scenario3.

water loss(million Yuan)

3.4 Scenario Definition
The primary water price is different in each activity as fig.4.
The primary price – also called benchmark – is comparedwith each scenario which is described in this section. Three
scenario conditions are summarized in table 2.
Scenario1 means that the amount of the water price charge forconstruction, mining, manufacture, electricity and others arelarger than the amount of price charge for agriculture, forestry,
livestock, fishery and agriculture service
Scenario2 means that the amount of the water price charge forconstruction, mining, manufacture, electricity and others areless than scenario1, and the amount of price charge for
agriculture, forestry, livestock, fishery and agriculture serviceis larger than sccenario1.
Then, water supply with water price reform doesn’t exceed
the present water authority income. The price that the watersupply exceeds the income of water authority is unfeasible.

change rate

OthersConstruction

Electricity

Manufacture

MiningFishery

Livestock

Forestry

Agriculture

quantity (million Yuan)

4.1 Welfare Loss with Quantity of Water Saving
Please start writing here. In this section, we compare thewelfare loss with quantity of water saving in each scenario.
We can analyze the welfare loss when the same amount of
0.0
0%water is saved in each scenario.
-0.5-1%
Fig.5 shows that the welfare loss with quantity of water-1.0
saving in each scenario. Scenario3 is the least of welfare loss
-2%
-1.5
when the equal amounts of water are saved. It shows that
scenario 3 is the best water price reform method. But-2.0-3%
scenario3 is also unfeasible, because water supply sector
Scenario1 Quantity
Scenario2 Quantity
needed to be subsidies when the water prices are under 0.7
Scenario1 Rate
Scenario2 Rate
(Yuan/m3) (dash line in fig.5), and the amount of water Fig.7 Change of quantity and change rate of final demand tosavings exceeds 80% of the amount of present water using save the water of 1 billion m3
488

-3%

Fishery

Agriculture sarvice

Scenario1 Quantity

Scenario2 Quantity

Scenario1 Rate

Scenario2Ratechange rate

-0.5

Others3%
Electricity

0.5Construction

9%

Mining

1.5

Manufacture

15%

Forestry

2.5

Livestock

21%

Agriculture

quantity (million Yuan)

3.5

Fig.8 Change of quantity and change rate of investment
demand to save the water of 1 billion m3

2

2%

-2

-2%

-6

-6%

Scenario1 Quantity-10%
Scenario2 Quantity

Scenario1 Rate

Scenario2 Rate

-10change rate

OthersConstruction

Electricity

Manufacture

MiningAgriculture sarvice

Fishery

Livestock

Forestry

Agriculture

quantity(million Yuan)

Looking at fig.7 and 8, the impact on final demand of
agriculture of scenario 1 is smaller than scenario 2, and theimpact on final demand of agriculture of scenario 1 is also
smaller than scenario2. Looking at fig.9, the impact on totalintermediate dmand of agriculture of scenario1 are larger thanscenario2. This means that the impact on domestic production
are caused by total intermediate demand (fig.10). Because,manufacture demand has a relative large amount ofintermediate from africulture , and water price reform of
scenario1, therefore there is a decrease of domestic
production in manufacture.
Fig.9 Change of quantity and change rate of total intermediatedemand to save the water of 1 billion m3

Fig.10 Input coefficient of manufacture
5. CONCLUSION
In this paper, we built a CGE model for water price reform.
0.3
0.25
0.2
0.15
0.1
0.05

6. ACKNOWLEDGMENTS

7. REFERENCES[1]

Alexander Smajgl, Romy Greiner, Colin Mayocchi, “Estimating theimplications of water reform for irrigators in a sugar growing region”,
Environmental Modeling and Software, vol. 21, 2006, pp. 1360-1367.
[2] Basanta K. Pradhan, M.R. Saluja, Shalabh K. Singh, Social Accounting
Matrix for India: SAGE Publication, 2006.
[3] Bell, C. and S, Devaraian, “Intertemporally Consistent Shadow Prices
in an Open Economy "Estimate for Cyprus"”, Journal of Public
Economics, Vol.32, 1987, p. 263-285[4] Bureau of Statistics of Zhangye, Zhangye Statistical Yearbook.
Zhangye, China. (in Chinese), 2007
[5] Chang K. Seung, Thomas R. Harris, MacDiarmid, “Economic impactsof surface water reallocation policies: A comparison of
supply-determined SAM and CGE models”, The Journal of Regional
Analysis and Policy. vol. 27, 1997, p.55-76.
[6] Frederick, K.D., “Balancing water demands with supplies: the role ofmanagement in a world of increasing scarcity”, World Bank TechnicalPaper No. 189, 1993.
[7] (GPBWR), Gansu Provincial Bureau of Water Resources, Gansu water
resource official report of 2002. Beijing : China Water Power Press. (inChinese), 2003.
[8] Gansu Statistical Bureau (GSB), Gansu Yearbook. Beijing(1979–2005): China Statistics Press. (in Chinese), 1978–2004.
[9] Gao, Q. Z. and Li, F. X. “Case Study of Rational Development andUtilization of Water Resources in the Heihe River Basin. Lanzhou”,
Gansu Science and Technology Press, 1991.
[10] Hanemann, M., Pricing as a tool for demand management: Workshop
water for the sity: strategic planning, demand management and network
losses control. Athens : National Technical University of Athens,
University of the Aegean, Water Supply and Sewerage Company ofAthens, 2001.
[11] Peter Berck, Sherman Robinson, George Goldman, “The Use ofComputable General Equilibrium Models to Assess Water Policies”,
California Agricultural Experiment Station Giannini Foundation ofAgricultural Economics. Working Paper No. 545. 1990.

Other

Construction

Electricity

Manufacture

MiningFishery

Agriculture
sarvice

Livestock

Forestry

Agriculture0

We applied the CGE model to the Zhangye economy. The
proposed method provided three water price reform scenarios
and these impacts on economy. Constant price for all
activities of welfare loss is the least when the equal amountsof water are saved. But it is also unfeasible, because the watersupply sector are needed to be subsidies when the water
prices are under 0.7 (Yuan/m3). And the amount of watersavings has exceed 80% beyond the amount of present waterusing when the water prices are more than 0.7 (Yuan/m3).
Looking at the welfare loss with quantity of water savings,
constant additional price charge for all activities has larger
decrease of welfare loss than constant rate price charge. But,
additional price charge for all activities affects mostly
agriculture. On the other hand, constant rate price charge for
all activities more decrease welfare loss than constant
additional price charge. But constant rate price charge for all
activities affects mostly agriculture and manufacture. In otherwords, constant rate price charge for all activities decreases
more in welfare loss, and decrease of domestic production are
incurred by mainly agriculture and manufacture. On the other
hand, constant additional price charge for all activities
decreases less in welfare loss and the decrease of domestic
production are incurred mainly by agriculture.

489

[12] Postel, S, Dividing the Waters: Food Security, Ecosystem Health andthe New Politics of Water Scarcity : World Watch Institute, 1996,
p.65-76.
[13] Takaaki Okuda and Takayuki Hatano, “A General equilibrium
modeling on water allocation in the Yellow River Basin,China”,
Proceedings Annual Meeting of Environmental Systems Research. 33,
2005, p. 65-72. (in Japanese)[14] Water Authority of Zhangye, Water Management Practices in Zhangye.
Zhangye, China. (in Chinese), 2007.
[15] Yan Chen, Dunqiang Zhang, Yangbo Sun, Xinai Liu, Nianzhong Wang,
Hubert H.G. Savenije, “Water demand management: A case study ofthe Heihe River Basin in China”, Physics and Chemistry of the Earth 30,
2005. p. 408–419.
[16] Yong Wang, Hong-lang Xiao, Rui-fang Wang, “Water Scarcity andWater Use in Economic Systems in Zhangye City. Northwestern
China” , Water Resour Manage 23, 2009, p. 2655–2668.
[17] Y. Wang, H.L. Xiao, M.F. Lu, “Analysis of water consumption use aregional input–output model: Model development and application to
Zhangye City. Zhangye” , Northwestern China Journal of Arid
Environments 73, 2009, p.894–900.

IW    IW  SW
W

(8)：saving of water authority

SW

Government Expenditure



G 
c

IG  SG  TG

G
c
DC

P

(9)

c

TG    IG
G

c



(10)

：quantity of government domestic consumer goodsG：share parameter of government demand：government saving (exogenous variable)

G
c

SG

Investment



I 
c

P

SH  SW  SG  IR

I
c

DC

(11)

c

SH    IH
H(12)

SW  IW    IW
W

Appendix. Model equation：quantity of investment demand for commodity：investment of rest of the world

 cI
H

：share parameter of household savingcIncome of Household

IH  P  L  P  K  TG  TR

(1)

TG    IG

(2)

L

K

H

H

G

：income of household

IH
P
PK
L
K
TG
TR

：wage
：capital rentL：labor endowment (exogenous variable)：capital endowment (exogenous variable)

H

H



(14)

V   DP

(16)

ca

V

V

a

a

H

W

a

0a

X

ca

：quantity of intermediate inputs：intermediate input coefficient

0a：value added input coefficient

Ca



DP

V

a

a



c

P
X

W

DC

V

a

a

a

 H  G  I

(17)

a

a

ca

c

c

(18)

c

：price of domestic consumer goodsDCa：direct income tax rate of household：direct tax rate of water authority

H

V

a

DP

DC   X：quantity of value added：value added tax rate

a
V

ca

c

：price of value added

a





 P X   1    P V
DC

P

a

P

(15)

Composite Commodity Market

(3)

a

V

a

：quantity of domestic productiona：government revenue

a

DP




Government Revenue
IG    P  V    IH    IWcaavariable)：share parameterIG：share parameter of investment demand

Domestic Production
X
X
X
X
X 
DPa  min 1a , 2 a ,  , ca ,  , Ca , a  ca Ca  0 a 
 1a  2 a
X   DP：income transfer to household from government：income transfer to household from the rest of the world(exogenous

G

(13)

I
IR

：quantity of Intermediate inputs
DPa ：quantity of domestic production
：production tax rate
ca

V

a

Household Expenditure

U H
c

 cH

(4)

c

c

 P  H   1     IH  SH
DC

H

c

c

Factor Market

K K  K

(5)

a

c

H

(19)

W

a

H 
c


P

H
c
DC

1     IH  SH 
HL  L

(6)

a

 
H

c

K

P H
0c

0c

 P  H 
DC

(7)

0c

c

c

H
P

c
DC

c



H
c

SH
H
P

0c

DC

0c

：capital endowment of enterprise (exogenous variable)

W

DC

0c

U

(20)

H

a

c







Interregional Balance
 E P  IR  TR   M P：utility level：quantity of household domestic consumer goods

DC

c

c

M
P

c

M

c

：price of domestic consumer goods：share parameter of household demand：household saving：initial quantity of household domestic consumer goods：initial price of domestic consumer goods

Water Supply Authority Expenditure

490

c

c

c

c

M



：quantity of import：import price (exogenous variable)(21)

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Post Earthquake Rapid Inspection Planning for Bridges in Western
Kentucky
Norimitsu Koike, Department of Civil Engineering, Aichi Institute of Technology
Issam E. Harik, Department of Civil Engineering, University of Kentucky

ABSTRACT: This paper presents a methodology for the rapid
assessment of bridge’s safety and functionality following a major
seismic event in flat rural areas. The proposed model considers a
combination of the bridge ranking prior to the seismic event and acomputer assist program to identify the “efficient completion time”
strategy for the post earthquake bridge inspection team to inspect thecritical bridge’s components damage level. Application of the
proposed methodology is conducted on the Purchase Parkway inWestern Kentucky. By setting the required inspection time to 40
minutes per bridge, it is estimated herein that a non trained
inspection team and without computer assist would conduct theinspection for all bridges 61 in a period exceeding 40 hours. On theother hand, by setting the time to 20 minutes for a well trained teamwith computer assist, it will need 22 hours to inspect all bridge 61,
and 5 hours to inspect provisionally the high risk bridges 9.

Keywords: Bridge rank, Inspection Strategies,
Post-earthquake inspection, Risk Potential
1. INTRODUCTION
Bridges have historically presented significant
vulnerabilities during major seismic events. Althoughengineers have taken steps to restore and reinforce bridges
against earthquakes, it is inevitable that unexpected damage
caused by lack of maintenance, poor design, etc., will occur.
Therefore, immediate dispatch for bridge inspection is very
important following an earthquake.
To assist in decision making and timely response to ensure
safety of the public and restore the transportation system
functionality, ShakeCast has been developed and provided
freely by U.S.Geological Survey (USGS) for emergency
responders[1]. In addition, Caltrans (California Departmentof Transportation) has developed and implemented a
Caltrans-specific version of ShakeCast, a post-earthquake
analysis tool for rapid evaluation of potential damages
structures[2].
In Japan, postearthquake response systems for road
networks have already been developed. The system iscomposed with early earthquake alerts which warn about
severe earthquake tremors a few seconds before they happen,a GIS database, ground shaking sensors, CCTV monitors andinspection tours by construction office staff[3].
Those post-earthquake response systems above mentioned
may be useful in the construction of real time information
facilities such as monitors, seismic sensors andcommunication lines. Although accurate real-time
information may be helpful for an efficient inspection tour, it
is difficult to extend those systems for a rural area road
network. In rural areas, the cost benefit balance of real-time

information systems may be lower than one in an urban area.
The financial demands for the construction and maintenanceof these systems are unavoidable. Consequently, forinspections that take place without real-time information
systems, a rapid inspection methodology is very important.
In Japan, post earthquake inspection systems for building
damage are conducted by architects and building engineers
who volunteer to evaluate the building damage and risk to the
public. The public sector organizes seminars to train
inspectors and registers them as volunteers. The results ofinspections, containing comments and remarks from resident
and visiting inspectors, are displayed on placards[4].
Following an earthquake in Japan, color inspection stickers
are seen at the site of a damaged building. However, thissystem is not used for bridge structures.
For bridge structures in Kentucky, the rapid inspection
system assists in providing a uniform approach for rating thedamage that has taken place[5]. The inspection teams can fillout an electronic form for all components of the bridge
structure with assistance from sample pictures and key pointcomments. A computer assist program provides
recommendation based on whether the observed damage wasnone, minor, moderate, or severe for each bridge component.
The recommendation is made based on the level of damage inthe bridge components. The inspection system interface in
Kentucky Transportation Center is composed of a laptop PC
that contains all the bridge data in Kentucky and a GPS unit
attached to the laptop that identifies the bridge once theinspector is at the site. The inspection systemin Kentucky is
inexpensive and does not require any special tools or
equipment.The system in Kentucky can be used not only byprofessional but also by a variety of personnel since it has thetools that assist the inspector embedded in the software. Thisis an important issue in the limitation of manpower andresource at a disaster.
This study attempts to understand the efficacy of theinspection system with a combination of pre-earthquake
investigations and rapid inspections. In addition, scenariosand discussions are presented for the inspection tour planningfor safety and road link recovery. The framework of theinspection system is discussed herein thorough an application
example using the Purchace Parkway which has one of thepriority routes on the Western Kentucky Parkway System.
2.

METHODOLOGY

2.1Outline of proposed method
The proposed time outline for the rapid inspection planningis shown in Fig.1. Firstly, pre-earthquake investigations

491

should be carried out to ascertain the present condition ofbridges. The data from the pre-earthquake investigation isused for forecasting possible damage and ranking the risk.
Rapid inspection immediately following an earthquake will
be carried out to make recommendations for maintaining
public safety and relieving traffic congestion. If theinspection staff has the data from pre-earthquake
investigations, their inspection is more effective and rapid.
They can check high risk bridges, identified in the
pre-earthquake investigation, as a priority.
When making recommendations for traffic relief, inspections
should be finished in a short time, and the shorter the betterfor rescue activities. However, a few hours may be requiredfor decision making to dispatch the inspection team(s)following the announcement of the magnitude and location of
the earthquake.
The objective of this study is to produce a recommendation
for a rapid inspection of bridges on priority routes. The areasof interest for this study are underlined in Fig.1.
2.2 Bridge risk ranking
The bridge risk ranking against an earthquake is needed tomake priorities for rapid inspection. In general, the risk
defines the product of the damage probability by the expected
loss. The damage to roadways may vary from minor damage
such as slight cracking of road surfaces to the bridge collapse.
The expected loss depends on the demand of the road. The
demand of the main roads is higher than that of the local roads
because the main roads may be used as the evacuation routefor residents and as designated routes of the relief teams.
Therefore, damage and demand have a direct influence on therisk potential. Consideration of demand for roads which are indifferent regions is also important. However, a uniform
demand can be on road sections that are direct routes. If roadsections are awarded a uniform demand, it means that therelative risk potential may be decided only by the damage
probability. Since the objective is to create a bridge risk
ranking, a relative calculation of the risk potential issufficient.
The challenge is how to calculate the damage probability.

The seismic retrofitting manual for preliminary seismic
evaluation can be used [6]. The manual describes the rating
method for a bridge retrofitting program. Since this manualaddresses multiple factors such as structural vulnerabilities,
seismic and geotechnical hazards, and bridge importance, it isa suitable method to define the post-earthquake inspection
ranking. The outline of the method is as follows:Firstly, the acceleration and importance coefficient are
determined. The acceleration is provided by any seismic data.
Some patterns of the importance of the bridge are described in
the original manual. Since all bridges on the parkways inWestern Kentucky are on essential or “priority” roads, theimportance coefficient of bridges targeted for application of
the proposed methodology can be classified as “essential”.
The seismic performance category (SPC) is determined by
the acceleration and the importance confidence by a tablecontained in the manual. If the bridge is categorized as SPC
A, this is omitted from the inventory. Secondly, the soil
profile type (S) and structural vulnerability rating (V) are set.
In addition, the seismic hazard rating (E) is modified by thesoil profile coefficient (S). Finally, the bridge rank (R) is
calculated based on a structural vulnerability rating (V) and aseismic hazard rating (E) as follows:
R=V*E

(1)

In general, a bridge of high R value is more vulnerable than
one with a low R value. Since V and E each range from 0 to
10, the value of R will vary from 0 to 100. The R value is easy
to understand as a relative index for the vulnerability of thebridge.
2.3 Rapid Inspection system
An essential tool is the inspection assist system, which has
been developed by the Kentucky Transportation Center for
rapid inspection after an earthquake. Although the target
region of the original system development is for the state of
Kentucky, if the bridge database of the target area is compiled
and installed into the system, its use in any region is possible.

Investigation of seismic zoom
Pre-earthquake investigation of bridges
Thebridge risk ranking and inspection priority
Pre-earthquake

EQ

Emergency Phase

The magnitude and location is announced
Decisionto dispatch therapid inspection tour
Departure the rapid inspection teamDisplay Recommendation for high risk bridges. The link is opened
Post earthquake inspection is done
Repair and Recovery
Fig.1. Time outline for rapid inspection planning
492

TIME

The rapid inspection procedures with the system are as
follows:
Firstly, the inspector checks the bridge type. The systemprovides the 17 types of bridge structure. If the database has
already been compiled, the inspector needs to select thebridge code number only. Secondly, the inspector makes avisual inspection. The points of view are embankment, span,deck, abutment, bearing, pier or column, and so on. An
internal program will provide an associated recommendation
based on whether the observed damage was none, minor,
moderate or severe for each component. Thirdly, therecommendation is made through the additional points fromeach component damage score. If the column leans, therecommendation is just “sever”. If some silly damage can see,the recommendation will be given by adding damage score.
The recommendation may be one of five grades- Green, Blue,
Yellow, Orange, and Red. Green means “Bridge Open”, Blue
means “Travel with Caution”, Yellow means “Reduced
Speed Limit”, Orange means “Bridge closed for Traffic
Except for Emergency Vehicles at Reduced Speeds” and Red
means “Bridge Closed completely”. The inspector prints out
the recommendation sheets for public awareness. Civil
engineers may also find the data useful information forfollow-up inspection.
If the time required for carrying out the inspection is too long,the efficacy of the system for rapid inspection is decreased, so
training and rehearsal drills for qualified personnel and rapid
inspection are required.
2.4 The strategy of an inspection tour
If a long time is required to inspect all bridges, theinspection team should limit the inspection to bridges withhigh priorities. However the risk potential of non-inspected
bridges cannot be ignored. Thus, the problem between therisk potential for inspection and the time required can bethought of as a trade-off relation. Normally, all bridges should
be inspected immediately following an earthquake. The
priority of inspections should be considered carefully. For the
strategy of the inspection tour, a check should be made to
determine whether the inspection tour can be completed
within the time limit for relief activity. If inspection for all
bridge is impossible, risk potential can’t be decreased. This
proposal’s methodology defines the sum of the R values asprevious risk potential in the region. After inspections andrecommendations for each bridge, the risk potential will
gradually decrease. When the inspection tour is finished, thesum of non-inspected bridges’R value will be shown as
non-inspected bridges’ risk potential. The non-inspected
bridges’ risk potential can be used as an important index forevaluating a post earthquake bridge inspection strategy.
The time required per bridge is an important parameter. Thisdepends on the proficiency of the system and bridge structure.
Of course the inspection can be carried out without thesystem; the team needs the handbook and calculation formaking the recommendation level by themselves. The team
may take a long time. When considering the time required forinspecting and assessing each bridge, this is ample motivation
for the training of personnel.
3

CASE STUDY

3.1 Example Area
The study area is in the Western Kentucky region located in
the New Madrid and the Wabash Valley Seismic zones. Thisis potentially one of the most destructive fault zones in the
United States. The broad impact has direct implications forvirtually every aspect of mutual emergency response.
Kentucky’s parkway were constructed during the 1960s and‘70s to augment the state’s interstate highways. Unlike roads
called parkways in other states, Kentucky’s parkways are notclosed to commercial and recreational traffic. Hence, it isessential that the parkways remain functional and operational
following an earthquake, especially, in the Western Kentucky
region.

Fig.2.Purchace Parkways in Western Kentucky and PGA for
a 250-year earthquake event8)

I.E. Harik et al. have evaluated bridges on and/or over the
parkways in Western Kentucky for seismic events[7], [8].
Their investigation included not only bridges, but also theircorresponding approaches and embankments. As a result of
their research, the risk to the Julian M. Carroll Purchase
Parkway (after this “Purchase PKWY”) is higher than otherparkways. On such purpose, I.E.Harik et.al.used the PeakGround Acceleration map showed in Fig.2 whichcorresponds to a 250-year earthquake event in Western
Kentucky. The 250-year earthquake event implies that there
is a 90% probability for the expected earthquake not being
exceeded in 250 years. PGA showed in Fig.2 were calculated
from a time history response spectra. The result of theinvestigation, preliminary analysis and ranking, shows that 13
bridges of the Purchase PKWY are deemed critical because
they were constructed in the 1960s, in which seismic design
was not taken into consideration. Purchase PKWY isconnected with the state of Tennessee, so it is an importantroute for response on multi-state issues.
3.2 Strategy and Scenario
One of the important issues of the inspection strategy is theaccessibility of inspection teams. However there are no deep
ravines and no elevated roads in the area. So, it is considered

493

that the inspection team may be able to move beyond thedamage bridge site. Therefore, the inspection teamtransportation speed may be set at about 48km/h(30miles/h).
There are two main purposes of the inspections. One is tomake a rapid recommendation of bridge safety. The other is toprovide road information as to whether people can travel fromorigin to destination smoothly. To accomplish the first
purpose, the team goes to bridges which are estimated to haveserious damage in advance, and displays a recommendation
regarding the risk to the public. Although people are able to
avoid dangerous bridges, they will be given more detailed
road link information later on. If the second purpose takes
precedence, the team checks all the bridges in the road section.
It is recommended to sweep the road link step by step.
However, it will require substantial time, as all bridges should
be inspected to ensure that they have only slight damage. It
will be difficult to accomplish the two objectives at the sametime due to the restriction of human inspection resources. The
problem is how to assign priorities and use inspection
manpower efficiently.
In order to provide both information of bridge safety androad information to the public timely, the dispatch patterns of
the Kentucky Transportation Cabinet personnel are defined
under the following strategies:
Strategy 1
The required time of inspection per bridge is varied between
20, 30 and 40 minutes. The time relates to the skill of theinspection team. This will have a large impact on theinspection completion time. In all cases, the inspection teamwill start from the I-24 junction.
Strategy 2
If the time required to inspect all bridges in one road section
is considered to be too long, a more efficient method may be
to omit the low risk bridges. The R value can be used forscreening in order to shorten the time for inspection in thisroad section. Since all of bridges are not inspected, it isnecessary to show how many risk potential of bridges are
dispelled by the inspection teams. The inspection percentage(I) is defined as follows,IS* 100(%)
T(2)

Where S is sum of inspected bridge’s risk potential, and T is
the total of all bridges’ risk potential
Strategy 3
If two inspection teams can be assigned, deciding how to
divide the roles between them is important for the inspection
strategy. This study proposes how to share the role based on
theR value. One team is the express team that checks the highrisk bridges. This team aims to display recommendations fordangerous bridges as soon as possible. This may permit the
avoidance of traffic congestion and maintain public safety. To
establish safety in this road section, the other team checks all
bridges except those which have already been awarded
recommendations by the express team. The aim of thismethod is to share the bridges that have border R value among

both teams. Therefore, the R value can be varied for sharing
the teams’ roles and to create the optimal strategy.
3.3 Result of Bridge Risk Potential
Fig.3 shows the frequency distribution of the R value versus
the Number of Bridges or “Number”of the Purchase PKWY.
The original data source is provided by the Kentucky
Transportation Center report [9]. Although many bridges arevaried between R=0 to 45, 8 bridges are over 75. It means thatsome specific bridges have a high risk potential and many
bridges have a lower probability of being damaged by anearthquake. In addition, two peaks can be seen between R=0
to 45. One peak can be seen in the low level range from 0-10.
The second peak is in the range of 20 to 30. The “Ratio”
values in Fig. 3 defines the number of bridges with an R valueless than or equal to a specific value of R on the abscissa axis
over the total number of bridges.

Fig.3. The frequency distribution of the bridge risk potential

4

RESULTS AND DISCUSSION

4.1 Results of Each Strategy
Strategy 1
The fastest case (or 20 minutes per bridge) implies aninspection with a computer assist system by a well trained
team. The middle case (or 30 minutes per bridge) is withcomputer assistance, but by a non-trained team. The results
are shown in Table 1.In all cases, the inspection teams startfrom the I-24 junction. Cases with the computer assist system
may decrease the time of inspection drastically. The
improvement of inspection time per bridge is effective forshort inspection tours.
Strategy 2
In strategies 1, there are recommendations to check all
bridges in the road section. However, there is the possibility
that inspections will take considerable time the 61 bridges.
Therefore, the case for checking only high risk bridges isdebatable. Through examination of Fig.3, it is found that highrisk bridges are limited in Purchase PKWY. For rapid
recommendations, it may not be necessary to inspect all
bridges here.

494

Table 1. Results of the four post earthquake inspection strategies
Provisional inspection
Inspection
Case
duration for high risk duration for all
bridges(Hours)
bridges (Hours)

Strategy

Strategy2(One team)Strategy3(Twoteams)

Well trained team with computer assist

-

22.1

100.0

Non trained team with computer assist

-

32.2

100.0

Non trained tem without computer assist
Bridges with R> 20 are inspected
Bridges with R> 35 are inspected
One team inspects bridges with R> 20 andanother team inspects bridges with R <20
One team inspects bridges with R> 35 andanother team inspects bridges with R < 35

9.4
4.7

42.4
-

100.0
72.8
45.6

9.4

14.4

100.0

4.7

19.1

100.0

As the Kentucky Transportation Center, bridges having an R
>35, are listed as high risk bridges[7]. In addition, Fig. 3
shows that the mid range bridges’ risk potential varies from
R=20 to 30. Thus, the target bridges for inspection are
selected based on their R value.
Fig.4 shows the risk potential in cases when the inspection
team checks bridges based on their risk potential. In case thatthe team inspects bridges having R> 20, the inspection will
finish within 10 hours. The inspection percentage I isapproximately 72.8%. The number of bridges inspected is 23
out of 61. This number implies that 37.7 % of the bridges areinspected in this road section. In case bridges having R>35
are inspected, the inspection tour will finish in approximately
5 hours. The inspection percentage I is 45.6%, and thenumber of bridges that have been inspected is 9 out of 61 (or14.8% of the bridges in this road section).

the inspection after 10 hours. At this point in time, because
inspection of high risk bridges has been finished, people may
be able to use the road slowly and carefully. The other team
may finish about 17 hours later. Bridges can receive risk
recommendations and people can travel using the roadaccording to the recommendation displayed at each bridge.
100.0
Total

Risk potential(%)

Strategy１

Percentage of
Inspected
Bridges (%)

90.0

HR Team

80.0

LR Team

70.0
60.0
50.0
40.0
30.0

100.0
20.0

90.0

All

10.0

R<35

80.0

R<20

0.0

Risk Potential(%)

70.0

0.0

5.0

10.0

15.0

20.0

Time(hour)

60.0

Fig.5.The decrease of the non-inspected bridges’ risk
potential by the express team for high risk bridges (R> 20)
and the team for low risk bridges (R< 20)

50.0
40.0
30.0
20.0
10.0
0.0
0.0

5.0

10.0

15.0

20.0

25.0

Time(hours)

Fig.4.The decrease of the non-inspected bridges’ risk
potential for different R values
Strategy 3
Fig. 5 shows the result of the inspection of bridges having
R< 20 and R> 20. The express team for high risk bridges(shown “HR Team”) inspects 23 bridges having R> 20, andanother team for low risk bridges (shown “LR Team”)
inspects 38 bridges having R< 20. The HR team may finish

Fig.6 shows the result of the inspection of bridges having R<
35 and R> 35. The HR Team (R> 35) inspects only 9 bridges,
and the LR team inspects 52 bridges. The HR Team mayfinish the inspection after 5 hours. The LR Team may finish
about 17 hours later.
Comparing the cases when the cutoff values for Rare 20 and
35, the completion time of the comprehensive bridge
inspection in the case of the cutoff value is 20 may be fasterthan in the case when it is 35. This is because the number ofbridges inspected is balanced among two teams in the casewhen the cutoff value is 20. However, in the case when the
cutoff value is 35, the inspection of the check high risk
bridges is faster.
Decision making regarding the inspection strategy is difficult
without estimating the magnitude of the earthquake in the495region. For example, if some bridges were to collapse, theroad link cannot be opened early by rapid inspection. Rapid
inspection for high risk bridges may be required for the public
safety. On the other hand, if many bridges suffer damage
without collapse, all bridges should be checked early to open
the road link. In this scenario, when a severe earthquake
occurs, the cutoff value of R = 35 is a better to choice.
100.0
90.0

Total

80.0

HR Team
LR Team

Risk potential(%)

70.0
60.0
50.0
40.0
30.0
20.0
10.0
0.0
0.0

5.0

10.0

15.0

20.0

25.0

Time(hour)

6 ACKNOWLEDGMENT

Fig.6.The decrease of the non-inspected bridges’ risk
potential by the express team for high risk bridges (R> 35)
and the team for low risk bridges (R< 35)
4.2 Discussion and Adoptable Strategy
Table 1 presents the result of all four strategies, and thefollowing can be deduced:
1. If an inspection team, without training, carries out theinspection for all bridges, an inspection time of over 40 hours
may be required. This may prove to be too late for relief
activity. By deploying a well trained team with an assist
system, the inspection will be finished in about 22 hours.
Training is significant in reducing the inspection time perbridge.
2. In cases that only high risk bridges are inspected by a welltrained team, the inspection will be finished in about 5 hours.
Although all bridges may not be able to receive
recommendations, people can go through the Purchase
PKWY carefully. This may be the swiftest strategy,
3. If two inspection teams are deployed, the high risk bridges
can be inspected in about 5 hours, and all bridges within 15
hours.
In conclusion, a well trained inspection team dispatched toinspect high risk bridges with R> 35, travelers can expect togo through the Purchace PKWY within 4.7 hours following amajor earthquake. It’s the adoptable plan for this area.

5

proposed model is inexpensive, as daily system maintenanceis not required. It may be a useful method for rural areas inwhich it is difficult to set many telemeter sensors. A sample
application for the Purchase PKWY in Western Kentucky
region was provided with the recommendation that inspection
time will be reduced dramatically by using more than one
team with good training and bridge risk ranking information.
As Western Kentucky is a flat land area, accessibility to thebridges’ sites by the inspection team is not accounted for inthis paper. We have fixed the speed of the inspection teamand haven’t done a sensitive analysis for this parameter. In
cases where ravines and steep cliffs form the landscape of thepriority route(s), urban areas, and other characteristics,
transportation of the inspection team(s) by helicopter or otherspecial transportation should be considered. Other strategies
will need to be formulated to suit the geographicalcharacteristics of each area.
The previous field survey was done in 2002 in Western
Kentucky area. And the R value was calculated by Seismic
retrofitting manual in 1995 version. However, new manual ispublished in 2006[10]. Equation (1) hasn’t changed in newversion. The attempt with 2006 version is need in future.

CONCLUTION AND FUTURE DIRECTIONSThis paper presented a rapid inspection tour planningfollowing an earthquake. The combination of the bridge
ranking with pre-earthquake investigation and the computerassist system is our principal proposal in this paper. The

We wish to acknowledge Kentucky Transportation Centerand University of Kentucky for their contribution.
7 REFERENCES[1]David Wald, Kuo-Wan Lin, Keith Porter, Loren L. Turner, “ShakeCast:
Automating and Improving the Use of ShakeMap for Post-Earthquake
Decision Making and Response”, Earthquake Spectra, Vol.24, No.2,
2008, pp.533-553.
[2] Loren L. Turner, David Wald, Kuo-Wan Lin; Shakecast Caltrans
Deploys a Tool for Rapid Post-earthquake Response, TR NEWS,No.261, pp.40-41. 2009.
[3] Yasuhiro Shoji,” Developing Earthquake Damage Detection andInformation Sharing Tools for Road Administrators”, Society for social
management systems 2009, P.81.
[4] The Japan Building Disaster Prevention Association,” Post-Earthquake
Temporary
Risk
EvaluationofDamagedBuildings”,
http://www.kenchiku-bosai.or.jp/english/file/epanfall.PDF, 2009.
[5] A.G. Sardo, T.E. Sardo, I.E. Harik,” Post Earthquake Investigation
Field Manual for the State of Kentucky, Kentucky, Kentucky
Transportation
Center”,
UniversityofKentucky,
2006,KTC-06-30/SPR234-01-1F.
[6] Buckle, I.G., and Friedland, I.M. ,“Seismic retrofitting manual for
highway bridges”, Federal Highway Administration, US Department ofTransportation. Publication, 1995,No.FHWA-RD-94-052.
[7] Issam E. Harik et.al,” Seismic evaluation of bridges on and over the
parkways in Western Kentucky Summary Report”, Kentucky
Transportation
Center,UniversityofKentucky,
2008,
KTC-07-02/SPR246-02-1F.
[8] R. Street, Z. Wang, I. E. Harik and D. Allen; Source zones, Recurrence
rates, and time histories for earthquake affecting, Kentucky, Kentucky
Transportation Center, University of Kentucky, KTC-96-4, 1996.
[9] JianXie, Issam E. Harik, Tong Zhao and Jindong Hu,“Preliminary
Seismic and Ranking of Bridges On and Over the Parkways in Western
Kentucky”, Kentucky Transportation Center, University of Kentucky,
2008, KTC-07-04/SPR246-02-3F.
[10] Ian Buckle, Ian Friedland, John Mander, Geoffrey Martin, Richard
Nuttand Maurice Power,” Seismic retrofitting manual for highway
bridges, Federal Highway Administration”, US Department ofTransportation. Publication,2006,No.FHWA-HRT-06-032.

496

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Organophosphoric Acid Triesters from Landfill Sites and Sewage Plants in Japan
Haruki Shimazu, School of Science & Engineering, Kinki University



ABSTRACT: Seven organophosphoric acid triesters (OPEs)
were measured in the water emissions from five active landfill sites
and three closed inactive landfill sites. Many kinds of OPEs weredetected and the highest concentration levels are 10,000ng/L ordersof magnitude. The OPE concentration levels for the closed inactive
landfill sites were almost same with those for the active landfill sites.
It will be necessary for the closed inactive landfill sites to docontinuous monitoring and further maintenance. The same OPEs
were measured in the inlet waters and the outlet waters from two
sewage plants. The OPE concentration levels in the inlet waters were
almost same with those for landfill sites. The OPE concentrations in
the outlet waters were lower than those in the inlet waters. Some
OPEs could be removed in the treatment process of sewage plants. It
is probably effective for OPE reduction to introduce the treatment
process into landfill sites.

Keywords: Organophophoric acid triesters, active landfill site,
closed inactive landfill site, sewage plant
1. INTRODUCTION
Organophosphoric acid triesters (OPEs) are ubiquitousenvironmental pollutants because OPEs have been used forplasticizer, flame retarder, antifoaming agent, surface active
agent and so on. Several OPEs have carcinogenic and/or
neurotoxic properties. Some reports indicate the elution fromplastic products and the emissions from landfill sites.
However, little is known about the occurrence and theemission characteristics of OPEs from landfill sites andsewage plants. The purposes of this paper are to measure
OPEs from active landfill sites, closed inactive landfill sites,and sewage plants and to investigate the OPE emission
characteristics from those sites and plants. It is determinedthat the water emissions from landfill sites and sewage plants
have an influence on OPE pollution in aquatic environments.
2. METHODS AND MATERIALS
2.1 Landfill site description
This study focused on five active landfill sites shown in Table
1. The landfill areas were from 8,469-100,816 m2.Many kindsof wastes were disposed in B landfill. Plastics were disposedin almost landfills.
Table 1. Active landfill sites monitored in this study
A
B
C
D
1995～

1991～

Landfill classification

Inert landfill

Controlled landfill

Area [m2]13,42164,688100,816

11,478

8,469

Capacity [m3]135,404

874,850

1,302,908

104,270

35,098Composition of waste

1990～

1989～

E

Opetation time

Inert landfill Inert landfill Inert landfill

Glass,

Ash, Glass,

Glass,

Glass,

M etals,

Incineration residue,

M etals,

Plastics,
Rubble.

Plastics,

M etals, Papers,

Plastics,

Rubber.

Plastics, Rubber,

Rubble.

Slag, Sludge,
Textiles.

1982～

Rubble.

In 2007, the water emissions from A landfill site, B landfillsite and C landfill site were taken three times on 11 November,
1 and 26 December. Furthermore, the water emissions from A
landfill site were taken five times on 20 November 2007 andsix times on 6 January 2008 to investigate temporal change ofOPE concentrations.
In 2008, the water emissions from A landfill site and B
landfill site were taken three times on 29 July, 17 October and
21 November. The water emissions from C landfill site weretaken four times on 14 June, 29 July, 17 October and 21
November.
In 2009, the water emissions from A landfill site and B
landfill site were taken three times on 20 July, 25 August and
25 September. The water emissions from C landfill site weretaken five times on 20 June 18 July, 25 August, 25 September
and 31 October. Furthermore, the water emissions from D
landfill site and E landfill site were taken one time on 25
August and 31 October, respectively.
This study focused on three closed inactive landfill sites
shown in Table 2. Many kinds of wastes were disposed in G
landfill site.
Table 2. Closed inactive landfill sites monitored in this study
F
G
H
Opetation time

1986～approx.1990

Landfill classification

Inert landfill

Area [m2]12,445

Capacity [m3]65,822Composition of waste

1986～2007
Controlled landfill

Inert landfill

20,445309,380-

Plastics,

Ash, Glass,

M etals,

Incineration residue,

Glass,

M etals, Papers,

Rubber.

Plastics, Rubber,-Slag, Sludge,
Textiles.

In 2008, the leachates from F landfill site were taken twice on
17 October and 21 November.
In 2009, the leachates from F landfill were taken five times on
20 June, 18 July, 25 August, 25 September and 31 October.
The leachates from G landfill site were taken twice on 18 Julyand 31 October. Furthermore, the leachates from H landfillsite were taken four times on 18 July, 25 August, 25
September and 31 October.
For active landfill site, there were 20 samples, 9 samples,
12samples, 1 sample and 1 sample from A, B, C, D and E,
respectively. For closed inactive landfill site, there were 7
samples, 2 samples and 4 sample from F, G and H,
respectively.
2.2 Sewage plant description
This study focused on two sewage plants shown in Table 3.

497

Table 3. Sewage plants monitored in this studyI

three typical SIM mass, each OPE is quantified by the largest
SIM mass. The detection limits were calculated fromthreefold values of signal-noise ratio in the baseline ofchromatogram. The recoveries and the variation coefficientsfor solved OPEs in this analysis ranged from 70% to 120%and from 7% to 20%, respectively.

J

Opetation commencement

1986

1972

Service area [km2]

28.81

72.13

Service population

163,209651,737

Treatment capacity [m3/d]

75,000380,000

Treatment process

Anaerobic-anoxic-oxic process

Step aeration process

Table 5. Operation condition for GC/MS in this study
5975B inert XL E/CI MSD (Agilent Technologies)

The treatment capacities for I sewage plant and J sewage plant
were 75,000 m3/d and 380,000 m3/d, respectively. The
treatment processes were anaerobic-anoxic-oxic process andstep aeration process.
In 2009, the inlet water in I sewage plant was taken one timeon 19 October. The outlet waters in I sewage plant were taken
twice on 18 September and 19 October. Furthermore, the inlet
waters in J sewage plant were taken eight times on 29
September, 2 and 3 October to investigate temporal change ofOPE concentrations. The outlet water in J sewage plant was
taken one time on 29 September. There were 3 samples and 9
samples for I and J, respectively.
2.3 Analytical methods and instruments
Seven OPEs shown in Table 4 were measured. Those areobject chemicals because many kinds of studies have beenreported that they are fluently detected in aquatic and airborne
environments.

HP-5MS(30m×0.25mmf, 0.25μm)

Oven temperature

50℃(1.5min)→20℃/min→180℃→5℃/min→280℃(1min)

Injection temperature
Injection method
Carrier gas
Detector temperature
SIM mass

250℃
Splitless, 2μL
He
230℃
TBP : 99, 151, 211 TBXP : 125, 199, 227
TCEP : 249, 205, 251 TDCPP : 191, 209, 193
TEP : 155, 99, 127 TEHP : 99, 113, 211
TPP : 326, 325, 215

3 RESULTS AND DISCUSSION
3.1 OPEs from active landfill sites
Seven OPEs were measured in the water emissions from fiveactive landfill sites. The solved OPEs and PAHs in the water
emissions from the active landfill sites are shown Table 6.
Table 6. OPE concentrations in the water emissions from the active landfills
Solved OPEs
Suspended OPEs
[ng/L]*
DR**[ng/L]*DR**
TBP
337(49.0-2670)
39/43
61.4(9.22-1340)
43/43
TBXP
1250(49.2-6110)
26/43
755(456-834)
5/43
TCEP
926(47.4-3850)
31/43
252(8.12-713)
18/43
TDCPP
51.9(20.6-83.1)
2/43
204(41.3-366)
2/43
TEP
1400(43.1-20900)
40/43
18.4(4.32-415)
36/43
TEHP
31.0(3.9-58.1)
2/43
8.77(1.17-29.0)
18/43
TPP
58.7(15.0-329)
13/43
49.1(18.1-204)
5/43* Median(min-max)** Detection rate

Table 4. Measured OPEs in this study
Chemicals
CAS Number abbr. logKow
Tributyl phosphate000126-73-8 TBP
4.00
Tri-2-butoxyethyl phosphate
000078-51-3 TBXP
3.75
Tri-2-chloroyethyl phosphate
000115-96-8 TCEP
1.44
Tris (1,3-dichloroisopropyl) phosphate013674-87-8 TDCPP
3.65
Triethyl phosphate
000078-40-0 TEP
0.80
Tris (2-ethylhexyl) phosphate
000078-42-2 TEHP
4.23
Triphenyl phosphate
000115-86-6 TPP
4.59* These abbreviations are used in this study.
** logKow values are obtained from reference

In the preparation, the water sample is filtered with glass fiber
prefilter (MILLIPORE AP40, nominal pore size 0.7μm).
OPEs in the filtered water are given as the solved OPEs.
OPEs in the suspended matter on the glass fiber are given as
the suspended OPEs.
Each 1000 ml filtered sample is passed into the solid phaseextraction column (WATERS PS-2) at the rate of 20 ml/min.
The column is dried with air by means of sanction pump. The
chemicals are eluted by passing 5 ml of dichloromethane
through the column. The extract is concentrated to 0.1 ml
under a N2 flow. Hexane is added to the extract until 2 ml.
Each glass fiber prefilter sample is dried in a dark place
overnight and put into a vial with 15ml of dichloromethane.
The chemicals are extracted with ultrasonic waves. The
extract is filtered and concentrated to 0.1 ml under a N2 flow.
Hexane is added to the extract until 2 ml. The chemicals inthese extracts are determined with gas chromatograph withmass spectrometer (Agilent technologies, 5975B inert XL
E/CI MSD).
The operation condition for GC/MS (gas chromatograph withmass spectrometer) is shown in Table 5. After qualifying by

ColumnFor active landfill sites, the solved OPEs for the high detected
frequencies were TBP, TBXP, TCEP, and TEP. The
suspended OPEs for the high detected frequencies were TBP
and TEP. The highest concentrations of the solved OPEs andthe suspended OPEs were 20900 ng/L for TEP and 1340 ng/L
for TBP.
The total concentrations and the component ratios of OPEs inwater emissions from active landfill sites were shown Fig.1.
The concentrations and component ratios were different ineach landfill site. The total of solved OPEs and suspended
OPEs from B landfill site were lower than those from theother landfill sites because many kinds of wastes were
disposed. However, the OPE concentrations for A and C
landfill sites were high. It may be due to plastic wastes
because typical OPEs containing plastic products such as
TBXP, TCEP, and TEP were so high [1].
The temporal changes of OPEs in the water emissions from A
landfill site on 20 November 2007 were shown in Fig.2. TBP,
TBXP, TCEP, and TEP ranged from approximately 1000
ng/L to 2000 ng/L. The specific trends of OPE concentrations
were not observed. The trends of OPE concentrations on 6

498

Table 7. OPE concentrations in the water emissions from the closed inactive landfills
Solved OPEs
Suspended OPEs
[ng/L]*
DR**[ng/L]*DR**
TBP
1490(321-6870)
13/13
50.4(4.05-3780)
13/13
TBXP
628(210-4230)
8/13
6.21
1/13
TCEP
89.3
1/13
29.2
1/13
TDCPP
52.5(37.0-277)
7/13
ND
0/13
TEP
1360(30.9-4290)
13/13
15.8(2.05-503)
10/13
TEHP
3.00
1/13
6.59(3.36-17.8)
10/13
TPP
ND
0/13
ND
0/13
* Median(min-max)** Detection rate*** ND means Not detected.

Total of solved and suspended OPEs [ng/L]

7000
6000
5000
4000

TBP
TBXP

3000

TCEP

2000

TEP

1000

sites, the OPE concentration levels were high. It shows thatthe closed inactive landfill sites need continuous monitoring
and further maintenance.

Others0AB

3.3 Effects of water temperature and rainfall on OPEemissions from landfill sites
Effects of water temperature and rainfall on OPE emissions
from landfill sites were investigated. Figure 3 shows therelationship between OPE concentrations in the water
emissions from the landfill sites and water temperature. The
correlation coefficient for the active landfill sites is so low
because wastes are continuously disposed in the active
landfill sites. However, the relationship between OPEemissions and water temperature for the closed inactive
landfill sites is a little significant. The disposed wastes
containing OPEs were so closed for a long time that watertemperature could depend on the elution characteristics for
OPEs. Figure 4 shows the relationship between OPEconcentrations in the water emissions from the landfill sites
and rainfall. For both active landfill sites and closed inactive
landfill sites, the significant relationship between OPEemissions and rainfall were not observed. The quantity ofwater emissions is not related with total of rainfall for sevendays before sampling. It could be due to the complex
mechanism of rainfall elution from landfill sites.

C

Total of solved and suspended OPEs [ng/L]

Figure 1 Total concentrations and component ratios ofOPEs in the water emissions from the active landfill sites

3000

2000

1000

0
10:00 12:00 14:00 16:00 18:00
TBP

TBXP

TCEP

TEP

TEHP

TPP

TDCPP

Figure 2 Temporal chenges of OPEs in the water
emissions from A landfill site (Nov. 2007)

30000

3.2 OPEs from closed inactive landfill sites
Seven OPEs were measured in the water emissions from threeclosed inactive landfill sites. The solved OPEs and PAHs in
the water emissions from the closed inactive landfill sites areshown Table 7.
For closed inactive landfill sites, the solved OPEs for the highdetected frequencies TBP, TBXP, TDCPP, and TEP. The
suspended OPEs for the high detected frequencies were TBP,
TEP, and TEHP. The highest concentrations of the solved
OPEs and the suspended OPEs were 6870 ng/L for TBP and
3780 ng/L for TBP. The OPE concentration levels for theclosed inactive landfill sites were almost same with those forthe active landfill sites. In spite of the closed inactive landfill
499

25000

Total of OPEs [ng/L]

January 2008 were almost same. It probably indicates thatthese OPEs were gradually dissolved from plastic wastes
and/or contaminated soils and emitted to the environment [2].

20000
15000
10000
5000
0
0

10

20

30

Water temperature [℃]
Active landfill

Closed inactive landfill

Figure 3 Relationship between OPEs in the water
emissions from landfillsites and water temperature40

Table 8. OPE concentrations in the inlet waters for the sewage plants
Solved OPEs
Suspended OPEs
[ng/L]*
DR**[ng/L]*DR**
TBP
764(441-14,700)
9/9
159(40.3-842)
9/9
TBXP
6920(1530-36200)
9/9
140
1/9
TCEP
3150(1320-4500)
4/9
ND
0/9
TDCPP
190(106-404)
7/9
ND
0/9
TEP
129(86.3-357)
9/9
36.1
1/9
TEHP
76.4(28.7-115)
9/9
266(118-508)
9/9
TPP
79.4(61.2-108)
9/9
ND
0/9
* Median(min-max)** Detection rate*** ND means Not detected.

3.4 OPE partition characteristics in water emission from
active landfill sites
Suspended OPEs / solved OPEs partition coefficients (Ksw)
in water emissions from active landfill sites are shown Fig. 5.
The shown OPEs were detected both filtered water andsuspended matter. Each dot means the median value of Ksw.
Each Ksw range means from minimum value to maximum
value. TEP is the most hydrophilic and TPP is the mosthydrophobic, as shown in Table 4. The trend indicates that
logKsw would be higher for hydrophobic OPEs.
7
6
5
4

Table 9. OPE concentrations in the outlet waters for the sewage plants
Solved OPEs
Suspended OPEs
[ng/L]*
DR**[ng/L]*DR**
TBP
1050(133-2690)
3/3
25.0(5.60-71.0)
3/3
TBXP
871(564-1400)
3/3
50.7
1/3
TCEP
1810(1600-2170)
3/3
ND
0/3
TDCPP
109(106-138)
3/3
20.8
1/3
TEP
143(103-203)
3/3
5.64(2.14-9.14)
2/3
TEHP
5.30(1.89-8.72)
2/3
15.6(8.77-22.5)
2/3
TPP
16.5
1/3
ND
0/3
* Median(min-max)** Detection rate*** ND means Not detected.

ng/L for TBP. The temporal changes of OPEs in the inlet
waters from O sewage plant were investigated. The specifictrends of OPE concentrations were not observed as those for
A landfill site shown in Fig.2. In the outlet waters, the highest
concentrations of the solved OPEs and the suspended OPEs
were 2690 ng/L for TBP and 71.0 ng/L for TBP. The
concentration levels for TBXP, TEHP, and TPP in the outlet
waters were lower than those in the inlet waters. Some OPEs
could be removed in the treatment process of sewage plants.
4 CONCLUSION

3
2
1
0
TEP

TCEP

TBXP

TBP

TEHP

TPP

Figure 5 Median and range of Ksw for detected
OPEs both filtered water and suspended matter

3.5 OPEs from sewage plants
Seven OPEs were measured in the inlet waters and outlet
waters for two sewage plants. The solved OPEs and PAHs in
the inlet waters and the outlet waters for the sewage plants areshown Table 8 and Table 9.
In the inlet waters, the solved OPEs for the high detected
frequencies were TBP, TDCPP, TEP, TEHP, and TPP. The
suspended OPEs for the high detected frequencies were TBP
and TEHP. The highest concentrations of the solved OPEs
and the suspended OPEs were 36200 ng/L for TBXP and 842

Seven OPEs were measured in the water emissions from fiveactive landfill sites and three closed inactive landfill sites. The
same OPEs were measured in the inlet waters and the outlet
waters in two sewage plants. Many kinds of OPEs weredetected and the highest concentration levels are 10,000ng/L
orders of magnitude. The OPE concentration levels for theclosed inactive landfill sites were almost same with those forthe attractive landfill sites. It will be necessary for the closed
inactive landfill sites to do continuous monitoring and further
maintenance. The OPE concentrations in the outlet waters
were lower than those in the inlet waters for sewage plants.
Some OPEs could be removed in the treatment process ofsewage plants. It is probably effective for OPE reduction to
introduce the treatment process into landfill sites.
5 REFERENCES
[1]
[2]

500

Kagawa Environmental Research Center,
http://www.k-erc.pref.kanagawa.jp/kisnet/
Yasuhara A, “Elution of tris(2-chloroethyl) phosphate from waste
plastics into water”, J.Environ. Chem., Vol.6, No.1, 1996, pp.43-47

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Ground Water Pollution around the Oyachi–Heizu District of Yokkaichi City
Takemoto Y, Takahashi M, Awaya K, Ioka M and Guo P, Kayou -cho 1200, Yokkaichi, Mie, Japan 512-8512
Yokkaichi University, Faculty of Environmental and Information Sciences

ABSTRACT: Arsenic pollution in the ground water has been
Table. 1 Features of the survey sites

widely found in the Tomida district in northern Yokkaichi city as
shown in previous studies. The biggest illegal disposal site ofindustrial waste in midland of Japan was found in this area.
In order to identify the source of this pollution, we haveinvestigated the ground water pollution around this illegal landfillsite. In this study, sampling of well and the seepage water was
carried out at 13 observation sites, and the concentrations of thefollowing materials such as pH, EC, COD, T-N, T-P, D-Fe, D-Mn,
NH4-N, NO3-N, Cl, Mg, Ca and 1,4dioxan were analyzed. In order to
identify the origin of this water pollution (D-Fe, D-Mn and Arsenic)
and to examine relation to the illegal disposal place, the analysis of
the ion ingredient of the ground water was also investigated.
However, the ion composition of the water in the place which left
remote from this landfill site was clearly different from the analysisof the ion ingredient of the water of this disposal site. The results
indicated that the pollution (Arsenic, etc) found widely in the area
which left far from the disposal site is not directly related with theillegal disposal site.

Site

Features of theground water

St1

Leaching water

St2

Deep well for
irrigation
Artesian water fromwell
Leaching water

800m

200m

1 km

200m

1.5km

---

Deep well for
irrigation
Deep well for
irrigation
Deep well for
irrigation
Leaching water fromretaining wall
Leaching water, neardisposal site
Leaching water neardisposal site
Leaching water neardisposal site
Leaching water from
disposal site

2 km

200m

2 km

200m

2.5km

200m

500m

---

100m

---

100m

---

200m

---

0m

---

St3
St6
St7
St8
St9
S12

Keywords: Arsenic pollution, Illegal disposal site,
Ion composition of ground water

St51

1. INTRODUCTION

St52

Arsenic pollution in the ground water has been widely foundin the Tomida district in northern Yokkaichi city as shown inprevious studies which were carried out in 2008[2]. The
biggest illegal disposal site of industrial waste in midland ofJapan was found in this area, and many toxic substances (As,Pb and 1,4Dioxan) are detected in the ground water of thelandfill site[1]. In order to identify the relation of thispollution with the disposal, we have investigated the groundwater pollution around this illegal landfill site[4][5].

St54
St55

Distance fromthe boundaryof dump site
500m

Well
depth
---

pH: Glass-electrode,
EC: Platinum electrode,
COD: Acidic Potassium per manganese method,
TN: Ultra violet absorption method,
TP: Molybdenum Ammonite method,
NO3-N: Simple analysis (of Pack test),
NH4-N: Simple analysis (of Pack test),
NO2-N: Simple analysis (of Pack test),
C l: Simple analysis (of Pack test),
F-: Simple analysis (of Pack test),
1,4Dioxan: LC-MS,
As: Data of the previous report [2][3]2. FEATURES OF THE INVESTIGATION SITES
Investigation was carried out at 12 sites, and outline of the site
is shown in Table 1. It was found that the higher values thanthe Japan environmental standard of Arsenic were detected in
St3, St8, St12 in previous research[2].
The concentrations of EC,COD,CL values are very highlevels around this illegal landfill disposal site.

3. SURVEY DATE AND ANALYTICAL METHODS

4. RESULTS AND DISCUSSIONS

Survey and observation was carried out in 9 May, 2011 andparameters and analytical methods used in this survey were as
the following features;

Results are shown as the following Tables. In order toinvestigate the data, survey sites are divided in three groups,
one group is the site of near this landfill site (within 300m
from dumping site, St51, St52 , st54, St55, to be indicated as
N-group), another group is the sites remote from this landfillsite (over 1 km away, St3 and St6 to St9, to be indicated as501

R-group), and the other group (St1, St2, St12, to be indicatedas O-group) in Fig 1.

Table. 2 Results of the analysis ( No.1)

Yokkaichi-Higashi I.C.
Yokkaichi University
Heizu station( Sangi )

St3

St6

St２

St３

pH
EC mS/cm

6.8
0.44

6.5
0.42

7.9
0.23

COD mg/L

4.6

2.5

1.7

TN mg/L

1.2

1.9

0.78

TP mg/L

0.02

0.06

0.56

St12 F mg/L
As mg/L
St51
D-Fe mg/L

1
0.005

1
0.002

0.6
0.012

8

10

<0.05

D-Mn mg/L

1

1.5

<0.25

NH4 mg/l

0.5

---

0.5

NO2 mg/l

<0.02

<0.02

<0.02

NO3 mg/l<0.5

Cl mg/L

30

50

6Mg mg/L

20

15

0.5

Ca mg/L

60

30

2

1,4Dioxan
mg/L

0.019

0.009

<0.005

St52
St55

Toshibasemiconductor
factory

St１

-

St54

Dump
site

Parameter

St2

St1

St7
St8

<0.5

Table. 3 Results of the analysis ( No.2)
St9

Fig.1 Investigation sites around the biggest
illegal disposal site
4.1 Grouping of the Results
D-Fe, D-Mn were detected in many site independently in allgroups. In N-group, where is regarded as the place of highly
influenced by the disposal, high concentrations of EC, COD,
CL, Mg, Ca, 1,4 dioxan were detected, these contaminants are
regarded as reaching substance from dumping site. However,contrarily, in R-group, concentration of these contaminant
were low. In the previous research, high concentration of
Arsenic was found in St3, St12, St8, these sites belong toR-group or O-group. From these results, we considered that
the As contamination found widely in these area is notdirectly related with the illegal waste disposal.
4.2 Results of Analysis presented by the Tables
We show our analyzed results of the ground water pollution
in the following Tables.

502

Parameter

St６

St７

St８

pH
EC mS/cm

6.3
0.18

6.6
0.22

7.4
0.19

COD mg/L

0.7

0.9

1.5

TN mg/L

0.11

0.74

1.2

TP mg/L

0.08

0.08

1.5

F mg/LAs mg/L

1.3

1<0.001

0.8
0.014

D-Fe mg/L

3

10

0.8

D-Mn mg/L

<0.25

1

<0.25

NH4 mg/l

0.2

---

0.7

NO2 mg/l

<0.02

<0.02

<0.02

NO3 mg/l

<0.5

<0.5

<0.5

Cl mg/L

6

4

3

Mg mg/L

5

8

3

Ca mg/L

8

10

3

1,4Dioxan
mg/L<0.005<0.005<0.005-

Table. 4 Results of the analysis ( No.3)
Parameter

St９

St12

St51

pH

6.5

6.9

6.8

EC mS/cm

0.19

0.71

0.60

COD mg/L

0.1

16

16

TN mg/L

2.8

1.8

6.4

TP mg/L

0.01

0.18

0.63

F mg/LAs mg/L

0.7

0.8
0.020

0.8
<0.0005

D-Fe mg/L

0

10

1

D-Mn mg/L

<0.25

0.5

<0.25

NH4 mg/l
NO2 mg/l
NO3 mg/l
Cl mg/L
Mg mg/L
Ca mg/L
1,4Dioxan
mg/L--<0.02
10
7
10
10

1
<0.02
<0.5
0
30
80
<0.005

0.5
<0.02
<0.5
30
8
50
0.024

-

4.3 Results of Analysis presented by the Radar Charts
Results of analysis are presented by the graph of the radar
chart in the following. We select the concentrations of
EC,COD,Cl ,Mg,Ca as drawing with the radar charts.
Group of high concentration of calcium ion (Ca) is St51 andSt52.
The high dense calcium ion (Ca) pollution indicates thatwaste wood and gypsum wall board from houses andbuildings seem to have been filled in the underground of thesesites.

Table. 5 Results of the analysis ( No.4)
Parameter

St52

St54

St55

pH

6.5

7.3

7.9

EC mS/cm

0.94

2.4

3.9

COD mg/L

23

14

64

TN mg/L

3.3

2.0

86

TP mg/L

0.45

0.25

0.35

1.3

0.5

7.5

-

F mg/LAs mg/L<0.001

D-Fe mg/L

8

0.3

10

D-Mn mg/L

2

5

<0.25

NH4 mg/l

---

0.5

50

NO2 mg/l

<0.02

<0.02

1.5

NO3 mg/l

<0.5

<0.5

20

Cl mg/L

50

600

1000

Mg mg/L

20

30

150

Ca mg/L

50

30

70

1,4Dioxan
mg/L

---

0.097

0.23

Group of high concentration of Cl ion is St54 and St55.
High values of chloride ion (Cl) indicate the influence ofcontamination from the waste disposal field.
The high concentration of chloride ion in the groundwater,
soil water, and river water are often observed because
chloride ion (Cl) is nonreactive and not sorptive and has noredox or precipitation. Near the waste disposal site
groundwater with high ammonium ion (NH4) and chlorideion (Cl) usually dominates.

503

5 CONCLUSION
In order to identify the origin of the ground water pollution(D-Fe, D-Mn, As, P) in northern Yokkaichi city, investigation
has carried out. Ion composition of the ground water in theplace remote from landfill site was clearly different from theion ingredient of the water of disposal site. These results
indicate that the pollution in order to identify the origin of theground water pollution (of Fe, Mn, As) found widely in thearea which left from the disposal site is not directly relatedwith the illegal disposal site. So, it is necessary that we should
advance the research more and more.
There are the places with less ground water pollution in thelong range from the waste disposal field. For example,
St3,St6 and St9.

6 REFERENCES[1] Survey report on matters of the

[2]

[3]

[4]

[5]

504

Yokkaichi-Ohyachi-Heizu disposal site,
Risk Hyouka-Hyou 2nd version ,2011.
Takahashi M, Awaya K, Ioka M, Takemoto Y, “Groundwater and Soil pollution near Ohyachi-Heizu Waste
Disposal Site”, J. of EIS, Yokkaichi University ,Vol, 11,No.2, 2008, pp. 27-31.
Takahashi M, Awaya K, Ioka M, Takemoto Y, “Groundwater Pollution in Yokkaichishi Tomida area”,J of EIS,
Yokkaichi University, Vol, 12, No.2, 2009, pp.39-42.
Awaya K,”The build up of the illegal dump of industrial
waste in the Ohyachi-Heizu district, Yokkaichi city,
based on the information released by Mie Prefecture”, J.
of EIS, Yokkaichi University ,Vol, 11, No.2, 2008, pp.
113-136.(in Japanese)
Ioka M, Takahashi M, Awaya K, Takemoto
Y, ”Visualization of Elution Tests and Ground water
Pollution of the Disposal Field in Ohyachi-Heize,
Yokkaichi, based on the Report of Mie Prefecture”, J. of
EIS, Yokkaichi University ,Vol, 13, No.1, 2009, pp.
11-23.(in Japanese)First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Characteristics of Damage to Structures Induced by The Tsunami of The 2011
East Japan Mega Earthquake
Ö. Aydan, Tokai University, Ocean Research Institute, Shizuoka, Japan
H. Tano, Nihon University, Dept. of Civil Eng., Koriyama, JapanT. Iwatate, Tokyo Metropolitan University, Dept. of Civil. Eng., Tokyo, JapanABSTRACT: The East Japan Mega Earthquake with a moment
magnitude of 9.0 occurred off the shore of Tohoku Region of Japan.
Although the seismologist of Japan anticipated an earthquake with amagnitude of 7.5 in the source area, the magnitude of this earthquake
was much greater than their anticipation. Furthermore, theanticipated level of the tsunami induced by this earthquake also
exceeded the level anticipated by the seismologists of Japan.
Therefore, the damage caused by the tsunami was tremendous andkilled more than 28000 people due to the insufficiency of thecounter-measures. The authors first present the characteristics of thetsunami and then describe damage to buildings and various
structures. In the final part of the article, lessons from thisearthquake are described and some recommendations for tsunami
resilient structural design are put forward.

front into the land as illustrated in Figure 2. If the strained
overriding plate slip along the plate boundary, it rebounds to
its unstrained state. As a result, the upper surface (before)
configuration changes to “after” configuration. Thisconfiguration change causes ground uplift near the plate frontand subsidence of ground away from the plate front. As aresult, the sea front moves towards the land and causing
inundated areas.
This caused inundation problems
particularly in the settlements during the high tides and theerosion along the seashore.

Keywords: Tsunami, East Japan Earthquake, damage, lessons
1 INTRODUCTION
The East Japan Mega Earthquake with a moment magnitude
9.0 took place at 14:46 (JST) on March 11, 2011[1]. The
earthquake was a subduction plate-boundary earthquake andthe rupture area was 450 km long and 200 km wide. This
earthquake caused gigantic tsunami waves, which destroyed
many cities and towns along the shores of Tohoku and Kanto
Regions of Japan. The casualties caused by this tsunami
exceed 28,000 people. The tsunami destroyed and heavily
damaged buildings of various types, transportation facilities
and infrastructures. The authors visited an area from Orai to
Hitachi in Kanto region and Iwanuma to Rikuzentakata in
Tohoku region and made observations on the damage ofvarious structures (Figure 1).
In this paper, the authors first present the characteristics ofthe tsunami and then describe damage to buildings andvarious structures. In the final part of the article, lessons fromthis earthquake together with those from the Aceh earthquake
are described and some recommendations for tsunami
resilient structural design are put forward.
2 CHARACTERISTICS OF THE TSUNAMI
The earthquake was an offshore plate boundary earthquakewith a sense of thrust faulting. The estimated inclination andoffset of the fault were between 14-16 degrees and 25-30 m,
respectively. The seabed was uplifted by 5 m while the landsank-down by about 0.8 m with a horizontal movement of 4.6
m [2]. The crustal deformation caused the intrusion of the sea

Figure 1: Investigation routes and epicenter location

Figure 2: An illustration of mechanism for the inundation
problem along the seashore.
The Port Authority Research Institute (PARI) operates
DARTS along the shores of Japan [3]. Figure 3 shows
locations of DARTS and their records of tsunami waves. The
DARTS are approximately 20 km away from the shore andthe sea depth varies between 125 m and 204 m. The
maximum tsunami height by DARTS was about 7m.

505

the appropriate warnings could not be issued until 15:33 [1].
When it was issued, it was too late for the cities and towns,
which were already heavily hit by the tsunami waves.

Figure 3: Locations and measured tsunami heights by
DARTS along the Tohoku region (modified from
PARI[2]).
There are two different definitions for characterizing thetsunami height, namely, the tsunami height at the shoreline
and inundation-height or run-up height (Figure 4). As it waspreviously difficult to measure the tsunami height at shoreline,
the inundation tsunami-height obtained from traces oftsunami front on the land is commonly quoted as the tsunami
height in the past. This may have tremendous values
depending upon how it is measured. For example, thenewspapers reported a tsunami height of 38.9 m at Aneyoshi(Miyako) measured by Tokyo University of Marine Scienceand Technology, which probably includes the splash zone.
The maximum tsunami height at shoreline in this earthquake
was measured as 15.3 m at Onagawa town by the authors.
Figure 5 shows the measured tsunami height together withestimations from some empirical relations.

Figure 4: Illustration of different tsunami heights (modified
from PARI[2])The maximum tsunami waves arrived at major towns andcities such as Kamaishi, Rikuzentakata, Kesennuma, Minami
Sanriku, Ofunato and Onagawa about 30-40 minutes after theearthquake. Figure 6 shows a computational result for thearrival time estimation for Ishinomaki City. The arrival timesof the maximum tsunami waves at several locations areroughly equal to those estimated from the single source modelwith the use of seabed topography and epicenter location.
Japan has a warning system based on a database ofpreviously computed results for scenario offshore
earthquakes. This system performed well until thisearthquake. Once the epicenter and magnitude of theearthquake are determined, the system evaluates the expectedtsunami heights at shores and their arrival times. As themagnitude of this earthquake could not be accurately
determined for a considerable time following the earthquake,

Figure 5: Comparison of observation with empirical relations
for tsunami shoreline and run-up heights [4,5,6].

Figure 6: Estimation of the arrival time of tsunami wave at
Ishinomaki.
3 COUNTER MEASURES AGAINST TSUNAMICounter-measures against tsunami can be broadly classified
into two categories, specifically, software measures andhardwaremeasures.
Softwaremeasuresinvolve
recommending people to escape locations with higher
elevations and/or earthquake and tsunami resistant highbuildings and terraces. However, when higher ground is far, it
may be extremely hard for elderly or sick people, pregnant
women and small kids to escape higher grounds, which may
be called as horizontal evacuation. At such places, the verticalevacuation in the form of escaping to higher buildings,
terraces and towers becomes the most effective softwarecounter measure [6].
Hardware countermeasures are a) Constructing highbreakwaters, b) Constructing evacuation terraces; c) Tsunami
gates at the estuary of rivers, and d) Planting trees. Japan withrich experiences of tsunami damage has tried very hard to
implement both software and hardware counter-measures.
However, tsunami shore-height was generally assumed to beup to 5-6 m in hardware countermeasures so far. The actionsof tsunami waves can be impact, surge, drag, buoyant andhydrostatic forces. Tsunami-resistant structures must be
capable of resisting against these forces.
4

RESPONSES OF STRUCTURES AGAINSTTSUNAMI WAVES AND ASSOCIATED DAMAGE

The authors investigated an area from Orai to Hitachi in506

Kanto region and Iwanuma to Rikuzentakata in Tohoku
region. Figure 7 shows the images of Rikuzentakata City
before and after the tsunami as an example. There were many
hardware counter-measures implemented in RikuzentakataCity. These counter measures were the construction of
breakwaters along the river and seashore, tsunami gates at theestuary of small rivers and planting very dense pine trees.
When two images are compared with each other, it is easily
noticed that none of hardware counter-measures wereeffective. Tsunami waves overtopped breakwaters, thepine-tree forest and tsunami gates resulting in the inundationof the almost entire city. This observation clearly indicatedthat if the tsunami wave height at shoreline is greater than theheight of any of these hardware counter-measures, thecounter-measures become ineffective. Figure 8 shows someviews of counter-measures after the tsunami in Rikuzentakata.
In this section, responses of structures observed in various
locations against tsunami waves are described.

Therefore, such structures should not be built intsunami-prone areas. Although stone masonry buildings arerare in the tsunami affected areas, they are much moreresistant against tsunami waves (Figure 10). Steel-framed
structures also performed poorly (Figure 11). The infill panelsof the steel-framed structures were punctured by impact
forces of tsunami and debris, and steel frames bended orbuckled. Reinforced concrete (RC) structures or steel
reinforced concrete (SRC) structures performed much better
among all building types (Figure 12). Except damage to someof RC or SRC buildings observed in Onagawa town, almost
all of these type buildings were structurally sound and they
were the only remaining structures in all visited towns andcities.

Figure 8: Views of the state of counter-measures after thetsunami.

Figure 9: Views of damage to timber buildings.

Figure 7: Images of Rikuzentakata City before and after thetsunami (images from GSI [2]).
4.1 Responses of Timber, Masonry, and RC and SteelFramed Buildings

In Onagawa town, RC or SRC buildings were even damaged
and some of them were toppled and dragged for several ten
meters (Figure 13). It was interesting to note that one building,which had piles with a diameter of 30cm and length of 4 m,
was uplifted from its piled foundations and dragged for about
70 m. The back-analysis of damage to RC and SRC buildingsin Onagawa should be the starting point for thetsunami-resistant building design in years to come.

Timber buildings are the less-resistant building type among
all other types of buildings (Figure 9). As they are very light,
there are easily uplifted and carried away and crushed.
507

4.2 Responses of Railways and Railway Bridges
Damage to railways and railway bridges occurred along
railway lines running close to the seashore. The piers ofrailway bridges were toppled, rails were dragged andembankments were scoured (Figure 14). Furthermore, trains
were uplifted and dragged for a considerable distances.

Figure 10: Views of damage to masonry buildings.

Figure 14: Damage to railways and railway bridges
4.3 Responses of Roadways and Roadway Bridges
Fundamentally damage to roadways and roadway bridges
were similar to that of railways and railway bridges and it
occurred when roadways were running close to the seashore.
The bridge decks were uplifted and dragged andembankments were scoured (Figure 15).
Figure 11: Views of damage to steel-framed buildings.

Figure 12: Views of damage to RC buildings.

Figure 13: Views of damage to RC or SRC buildings.

Figure 15: Views of damage to roadways and roadway
bridges.
508

4.3 Responses of Airports
The Sendai Airport, which is about 1 km away from the sea
shore and at the elevation of 2m above sea, was inundated by
the tsunami. Tsunami height at the airport was about 4.9-5.7
m. The runway and ground floor facilities were all damagedby the inundation, mud and debris of the tsunami (Figure 16)
The airport was cleaned up and it started its function on April
13, 2011 with some domestic flights. The piers andembankments of railway and roadway bridges were also
scoured and at some locations and there were deck falls(Figure 17).

Figure 18: Views of souring in embankment of roadways andrailways in Sendai Airport
4.5 Responses of Rock Slopes and Tunnels
The damage to slopes by tsunami waves is due to drag andhydrostatic forces. In some places, shotcrete was peeled off(Figure 19). Tsunami waves inundated the West Makiyama
Tunnel (Ishinomaki) up to 1 m. However, the damage to thetunnel itself was none (Figure 20).

Figure 16: Views of damage to Sendai Airport

Figure 19: Views of damage to slopes

Figure 17: Views of souring at piers and embankment damageof roadways and railways in Sendai Airport
4.4 Responses of Embankments
The damage to embankments by tsunami waves is due totsunami induced liquefaction and erosion from drag forces.
Such damages were observed along river embankments, atrailway and roadway embankments (Figure 18). Furthermore,
the erosion of approach embankments of railway androadway bridges was also common in the tsunami-affectedareas.

Figure 20: Views of inundated Makiyama Tunnel509

4.6 Responses of Gigantic Breakwaters and Causes of
Their Damage
Japan constructed gigantic breakwaters at various locations.
Breakwaters at Kamaishi, Ofunato, Rikuzentakata and Tarowere all damaged and they could not perform as they were
supposed to be. The Kamaishi gigantic breakwater was 61 mhigh from the sea bottom. Figure 21 shows a cross-section of
the gigantic breakwater. The caissons of the breakwater weresank down and tilted after the tsunami. The video images
taken during the arrival of tsunami clearly demonstrated thatthe tsunami surf appeared at locations where they should not
be. This simply implied that the damage to caissons of the
breakwater already took place before the arrival of thetsunami due to ground shaking induced by the earthquake. It
is most likely the rubble mound below the caissons sankdown due to ground liquefaction of seabed soil. This isactually a commonly observed phenomenon in the pastearthquakes as well as in this earthquake (Figure 22).

Figure 21: An illustration of cross section of Kamaishi port
gigantic breakwater [3].

2) Although Japan has been well prepared for tsunami and its
disasterous effects, the anticipated tsunami height for
implementing hardware measures was much less than thatinduced by this tsunami.
3) The observations on the damage induced by the tsunami of
the 2011 East Japan mega-earthquake were basically similar
to those of 2004 Aceh (or Off Sumatra).
4) Timber buildings can not stand against high tsunami waves.
Steel framed structures are also weak against tsunami waves
due to the weakness of the infill panels. Reinforced concretebuildings performance was best to resist gigantic tsunami
waves provided that they are well built against groundshaking. Therefore, timber buildings must not be allowed intsunami-vulnerable areas.
5) This earthquake also showed that the vertical evacuation in
relatively flat areas was important to save lives.
6) The effectiveness of planting trees against tsunami must be
re-evaluated.
7) Very strong impact, surge, buoyancy, and dragging forcesof the tsunami waves were the primary factors for the damage
to bridges despite they are well-designed against groundshaking.
8) The failure of breakwaters to protect the settlements may
have been caused by the failure of seabed ground due to thestrength loss resulting from the reduction of effective stressunder high prolonged tsunami waves and/or groundliquefaction.
9) The inundation of ground along sea shores is due torebound of strained overriding plate. It would be quitedifficult to preserve the previous shorelines for a considerable
period of time. The current shoreline may further retreat due
to erosion by high waves.

6

REFERENCES

[1]
[2]

Japan Meteorological Agency (JMA), Tokyo, Japan
Geographical Information Authority of Japan (GSI): Special web sitefor East Japan mega earthquake.http://www.gsi.go.jp/BOUSAI/Port Authority Research Institute.(PARI): Special web site for East
Japan mega earthquake.http://www.pari.go.jp/info/tohoku-eq/
Abe, K., “ Size of great earthquakes of 1837-1974 inferred fromtsunami data”, J. Geophys. Res. 84(B4), 1979, 1561-1568.
Iida,K, “Magnitude,energy,and generation mechanisms of tsunamis
and a catalogue of earthquakes associated with tsunamis”, inProceedings, Tsunami Meetings Associated with the Tenth Pacific
Science Congress, pp.7-18,Int.Union of Geod.and Geophys.,Paris,
1963.
Aydan, Ö., “Seismic and Tsunami Hazard Potentials in Indonesia witha special emphasis on Sumatra Island”. Journal of The School ofMarine Science and Technology, Tokai University, 2008, Vol.6, No.3,
pp.19-38.

[3]
[4]
[5]

Figure 22: Settlement of tetra-pods in Shin-Urayasu causedby the 2011 East Japan Earthquake.
5[6]

CONCLUSIONS

The following conclusions and lessons may be drawn fromthe observations reported in this article and previous
earthquakes:
1) The shore height (inundation height at seashore) is themost important parameter for the design of structures against
potential tsunamis. The shore height induced by this tsunami
is in accordance with the estimations from the empirical
relations by Abe (1979) and Aydan (2008).

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First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Characteristics of Strong Motions Induced by the 2011 East Japan Mega
Earthquake with an Emphasis on Tokyo Bay Area
T. Iwatate, Tokyo Metropolitan University, Dept. of Civil. Eng., Tokyo, JapanÖ. Aydan, Tokai University, Ocean Research Institute, Shizuoka, JapanABSTRACT: The 2011 East Japan Mega Earthquake with amoment magnitude 9.0 occurred on March 11, 2011. This
earthquake shook almost the entire Japan. There are several strong
motion networks and these networks recorded strong groundmotions, which may be of great value for the earthquake engineeringcommunity, as it is very rare to have such a rich source of strong
ground motions. The authors processed the acceleration records
obtained by NIED (namely, K-NET and KIK-NET), Port Authority
Research Institute (PARI) and Tokyo Metropolitan University inthis study. In this article, the authors present the outcomes of theirstudies on the strong motions induced by the East Japan megaearthquake with an emphasis on those in Tokyo Bay Areas.

Keywords: 2011, East Japan, earthquake, strong motion
1 INTRODUCTION
The 2011 East Japan Mega Earthquake with a moment
magnitude of 9.0 occurred off the shore of Tohoku Region ofJapan [1]. The earthquake was subduction plate-boundary
earthquake and the rupture area was 450 km long and 200 km
wide. Although the seismologist of Japan anticipated anearthquake with a magnitude of 7.5 in the source area, themagnitude of this earthquake was much greater than theiranticipation.
This earthquake shook almost the entire Japan. There areseveral strong motion networks operated by several
organizations and institutes. These networks recorded strong
ground motions, which may be of great value for theearthquake engineering community, as it is very rare to havesuch a rich source of strong ground motions. The authors
processed the acceleration records obtained by NIED[2](namely, KNET[3] and KIKNET[4]), Port Authority
Research Institute[5] (PARI) and Tokyo Metropolitan
University (TMU). This earthquake also induced heavy
ground liquefaction particularly in reclaimed ground aroundTokyo Bay area. As an inland earthquake with a magnitude of
7.2 is anticipated in Tokyo Bay area, the ground shaking
characteristics of Tokyo Bay area are of great importance inview of seismic responses of various structures. In this article,
the authors will presents the outcomes of their studies on thestrong motions induced by the East Japan mega earthquakewith an emphasis on those in Tokyo Bay Areas.

institutions. The Japan Meteorological Agency (JMA) had
difficulties to determine the magnitude of the earthquake
soon after the earthquake due to several technical reasons andthe magnitude was updated to 9 after several days.
The earthquake occurred along the subduction zone between
North American plate (NAM) and Pacific plate (PAC)(Figure 1) and the inclination of the rupture plane wasestimated to be 14-16 degrees. The faulting was due to thrustfaulting with an estimated offset of 25-30 m. The seabed was
uplifted by 5 m while the land sank-down by about 0.8 m witha horizontal movement of 4.6 m [6]. The rupture initiated in
the area where an earthquake with a magnitude of 7.5 wasanticipated and it propagated bi-laterally. The rupture plane isabout 450 km long and 200 km wide and it consisted of 4
distinct segments. There have been numerous aftershocks andsome of aftershocks were greater than magnitude 7 (Figure 2).
The aftershock activity occurred in and around the rupturearea and some of shocks along the boundary between North
American Plate and the Eurasian plate (EUR) took place in
Shizuoka, Nagano and Akita prefectures. The rupture front
stopped at the boundary between the Philippine Sea Plate(PHS) and Pacific Plate. As the stress distribution entirely
changed after this earthquake, it is feared that it may initiate
mega-earthquakes in Kanto, Tokai, Tonankai and Nankai
subduction zones.

2 CHARACTERISTICS OF THE 2011 EARTHQUAKEThe magnitude of the 2011 East Japan Mega-earthquake wasestimated to be between 8.9 and 9.1, depending upon the

Figure 1: Location of the 2011 East Japan earthquake andtectonics of the region (modified from GSI [6]).

511

(a) NS component

Figure 2: Aftershock distributions as of April 8, 2011.
3 STRONG-MOTION NETWORKS
There are several strong-motion networks in theearthquake-affected area, operated by several institutes andorganizations. Following the 1995 Kobe earthquake(Hyogo-ken Nanbu earthquake), Japan established two strong
motion networks, namely, KNET and KiKNET and they areoperated by the NIED. The KNET network is for measuringstrong ground motions at the the free-field ground surface
while the KiKNET network is for measuring thestrong-motions at the bed-rock and ground surface. The
Japan Meteorological Agency also has a strong-motion
network throughout Japan, which is used to estimate intensity
scale and its distribution. However, the strong-motion recordsof this network are not always easily accessible. The Port
Authority Research Institute (PARI) has a strong motion
network covering the major ports in Japan and their data isaccessible through the Internet. In addition, Tohoku
University, Electrical Power Companies, Japan Railways,
Japan Expressways (NEXCO), Building Research Institute(BRI), Tokyo University, Tokyo Metropolitan University andother institutes and organizations have their own networks.
However, their records are generally not accessible throughthe Internet. In this article, the records obtained by the KNET,KiKNET, PARI and Tokyo Metropolitan University (TMU)
are processed and the results are reported.

(b) EW component

4 CHARACTERISTICS OF STRONG-MOTIONS
Figure 3 shows the NS, EW and UD components ofacceleration records of some selected strong motions stationsof KNET and KiKNET networks from Shimizu to Hachinohe.
As noted from the figure, the records strongly reflect therupture propagation. While the strong motion stations to thenorth of the epicenter clearly shows the rupture sequence of 4
segments, the records of the stations to the south of theepicenter are superimposed so that it is indistinguishable.

(c) UD Component
Figure 3: Acceleration records of some selected strong
motion stations of KNET and KiKNET of NIED.
Figure 4 shows the maximum ground acceleration contours,
which are obtained using the data of KNET and KiKNET.
The highest ground acceleration reaching to the level of 3g
was recorded at the Tsukidate strong motion station of KNET.

512

Although the ground conditions are very good at this station,
it is still difficult to understand why such a high groundmotions recorded at the Tsukidate station.

L*  a  be cM w

(3)

The specific values of constants of Eqs. (1)-(3) for thisearthquake are given in Tables 1 and 2.
Table 1: Values of constants in Equation (2) for inter-plate
earthquakes
Amax
Vmax

A

B(m/s)

D

E

F

G(Mw)

2.8
0.4

1000
1000

0.5
0.5

1.5
1.5

0.5
0.5

1.05
1.16

Table 2: Values of constants in Equation (3) for earthquakes
Faulting Type

a

b

c

Normal faulting
Strike-slip faulting
Thrust faulting

30
20
20

0.002
0.002
0.002

1.35
1.40
1.27

Figure 4: Maximum acceleration contours.
The attenuation of maximum ground acceleration andmaximum ground velocity are shown in Figure 5 together
with several empirical attenuation relations. As noted fromthe figures, none of attenuation relations [7,8,9,10] canestimate the attenuation of maximum of ground accelerationand velocity except the relation proposed by Aydan and Ohta
[11]. In other words, so-called NGA(Next Generation
Attenuation)-like relations can not estimate the strong motion
attenuation recorded during the 2011 East Japan
Mega-earthquake.
Aydan and Ohta [11] relation given below in which theposition of observation point, inclination and length ofearthquake fault and ground properties are considered:(a) Attenuation of maximum ground acceleration(1)
 F1 (V ) * F2 ( R, ,  , L* ) * F3 ( M )s*where Vs,  ,  , L and M are the shear velocity of ground andmax

the angle of the location from the strike and dip of the fault(measured anti-clockwise with the consideration of the*mobile side of the fault) and earthquake magnitude. L (inkm) is a parameter related to the half of the fault length. The
following specific forms of functions in Equation (1) are as
follow(2a)
F1 (Vs )  Ae Vs / B

F2 ( R,  ,  , L* )  e  R (1 D sin   E sin  )(1 F cos  ) / L(2b)

F3 ( M )  e M / G  1

(2c)

2

*

*

L , which is the half of the fault length, is related to themoment magnitude in the following form(b) Attenuation of maximum ground velocity
Figure 5: Attenuation of maximum ground acceleration andvelocity.

513

spectra of acceleration and velocity are particularly high at aperiod less than 0.2s. As the damage was particularly light inthe close vicinity of the Tsukidate strong motion station, thismay be explained through the characteristics of the strong
motion spectra of acceleration and velocity although the
maximum values of strong motions were unusually high.

(a) Acceleration response spectra(b) Velocity response spectra
Figure 6: Response spectra of acceleration records at several
selected station nearby the earthquake epicenter
The acceleration and velocity response spectra of selected
strong motion stations in the close vicinity of the earthquake
epicenter are computed and shown in Figure 6. It is important
to notice that the response spectra of the Tsukidate strong
motion station is extremely high and such high values of
response spectra can not be enveloped by neither Japaneseseismic design code or other countries code. The response

Figure 7: Response spectra of acceleration records at Sendai
K-NET and PARI strong motion stations.
One of the interesting observations regarding the strong
motion records recorded by several institutes was that therecords were quite dissimilar, quantitatively. Figure 7 shows
the acceleration response spectra of records in Sendai City.
The strong motion stations of the KNET and PARI are very
close to each other. While very high motions are observed in

Figure 8: Spectral response accelerations of selected strong motion stations around Tokyo Bay.

514

the records by the KNET for a considerable period range, thehigh ground motions are only limited to the short period in therecords by the PARI. Furthermore, the maximum groundaccelerations of the PARI is almost twice or three-times
smaller than those of the KNET. Similar situations wereobserved at other strong motion stations. This topic deserves
very detailed further investigations including thecharacteristics of the strong motion instruments as well as
local ground conditions.

motions at Chiba strong motion station of KiKNET are
amplified for periods more than 2 seconds.

(a) K6-Zushi Elementary School

5 CHARACTERISTICS OF STRONG-MOTIONSAROUND TOKYO BAY AREA
There are many strong motion records obtained by various
institutes and organizations around Tokyo Bay area. In thisarticle, the strong motion records obtained by KNET, PARI
and Tokyo Metropolitan University (TMU) are only reported.
Tokyo Bay area is roughly 375 km away from the earthquake
epicenter, the surface ground accelerations ranged between
100 to 280 gals. Figure 8 shows the acceleration spectra ofstrong motion stations around the Tokyo Bay. The strong
motion stations such as Yokosuka and Shinjuku are locatedon relatively hard ground while the rest of stations are on thesoft sedimentary ground. Particularly, the spectral
amplification of ground motions with long period
components at Urayasu, Yokohama, Kisarazu strong motion
stations are of great significance.
Next, the spectral acceleration of surface ground motion
taken at Chiba by KNET, KiKNET and PARI strong motion
networks are compared with each other (Figure 9). As notedfrom the figure, the absolute values are quite different fromeach other although the results seem to be similar. Without
any doubt, the differences may be related to the difference ofthe ground conditions at each station. However, thedifference in the characteristics of the strong motion devices
might also have some influences on the measured strong
motions. This aspect still needs some further clarification as it
was pointed out in the previous section.

(b) Hiroba(c) Yokohama(d) Shinagawa

Figure 9: A comparison of acceleration response spectra at
Chiba of K-NET, KiK-NET and PARI strong
motion networks.
Figure 10 shows the acceleration spectra of strong motions
taken at ground surface and bedrock (underground) at K6
Zushi Elementary School (TMU), Hiroba Tokyo
Metropolitan University (TMU), Yokohama (KiKNET),
Shinagawa (PARI) and Chiba (PARI & KiKNET). As notedfrom the figure, The ground motions are highly amplified for
periods up to 1 second. However, the horizontal ground(e) Chiba
Figure 10: Acceleration response spectra of K6 Zushi
Elementary School, Hiroba Tokyo Metropolitan
University, Yokohama, Shinagawa and Chiba.
Except the response acceleration at Yokohama station of
KiKNET, the ground amplification generally ranges between

515

3-5 times. However this value reaches up to 20 times at
Yokohama strong motion station of KiKNET. This is anextremely high value and it deserves further detailed studies
before drawing any conclusion.
6 CONCLUSION
The following conclusions can be drawn from this study as
follow:
1) Numerous strong motions stations operated by several
institutes and organizations recorded ground motions inducedby the 2011 East Japan Mega-earthquake. It is probably the
first case to obtain such an extensive source of information
for a mega earthquake. Ground motions recorded by the
KNET and KiKNET networks of the NIED were extremely
high particularly in the vicinity of the initial rupture area.
2) The current attenuation relations for strong ground motions
are mostly incapable of estimating the attenuation of
maximum ground motions. However, the ground motion
records differed depending upon the network even though
they were very close to each other. This topic strongly
deserves very detailed further investigations including thecharacteristics of the strong motion instruments as well as
local ground conditions.
3) Although the Tokyo Bay area was roughly 375 km awayfrom the earthquake epicenter, the surface groundaccelerations ranged between 100 to 280 gals. The groundmotions were amplified at locations with thick soft
sedimentary deposits. Particularly, the spectral amplificationof ground motions with long period components at Urayasu,
Yokohama, Kisarazu strong motion stations were of greatsignificance.
4) The ground motions around the Tokyo Bay area werehighly amplified up to 1 second. However, the horizontal
ground motions at Chiba strong motion station of KiK-NET
were amplified more than 2 seconds.

[9]

Aydan, Ö. “Comparison of suitability of submerged tunnel and shieldtunnel for subsea passage of Bosphorus (in Turkish)”, GeologicalEngineering Journal, 25(1), 1-17, 2001.
[10] Ulusay, R., E. Tuncay, H. Sonmez and C. Gokceoglu “ An attenuation
relationship based on Turkish strong motion data and iso-acceleration
map of Turkey”, Engineering Geology, V. 74, 265-291, 2004.
[11] Aydan, Ö., Ohta, Y. “A new proposal for strong ground motion
estimations with the consideration of characteristics of earthquake
fault”, Seventh National Conference on Earthquake Engineering,
Istanbul, Turkey. (on CD), 2011.

ACKNOWLEDGEMENTS
The authors sincerely acknowledge Tokyo Metropolitan
University, National Research Institute for Earth Science andDisaster Prevention (NIED) and Port Authority ResearchInstitute (PARI) for making available the strong motion of the2011 East Japan Mega Earthquake through the Internet.
7

REFERENCES

[1]
[2]

Japan Meteorological Agency (JMA), Tokyo, Japan
National Research Institute for Earth Science and Disaster Prevention(NIED). http://www.hinet.bosai.go.jp/topics/
KNET: http://www.kyoshin.bosai.go.jp/kyoshin/
KiKNET: http://www.kik.bosai.go.jp/kik/Port Authority Research Institute.(PARI): Special web site for East
Japan mega earthquake.http://www.pari.go.jp/info/tohoku-eq/
Geographical Information Authority of Japan (GSI): Special web sitefor East Japan mega earthquake.http://www.gsi.go.jp/BOUSAI/
Joyner, W.B. and Boore, D.M. “Peak horizontal acceleration andvelocity from strong motion records from the 1979 Imperial Valley
California Earthquake”, Bull. Seis. Soc. Am., 71(6), 2011-2038, 1981.
Fukushima, Y., Tanaka, T. and Kataoka, S. " A new attenuation
relationship for peak ground acceleration derived from strong motion
accelerograms.,” 9th WCEE, Tokyo, 343-348, 1988.

[3]
[4]
[5]
[6]
[7]

[8]

516

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051


Inquiry of the Value of Parks in the Characteristics and Use of Park
through Urban Revival Planning Projects in Maebashi City
*Shinya TSUKADA, **Tetsuo MORITA and ***Akira YUZAWA

ABSTRACT: This study looks at Maebashi City in Gunma
Prefecture, one example of a local city with a stock of many parks
from post-war urban revival planning projects, and aims to consider
future park developments by evaluating park planning and use
through an understanding of (1) the features of park arrangement inurban revival planning projects, and (2) the changing use of plannedparks.
First, this study will demonstrate the features of park arrangement
planning in post-war urban revival planning projects. Next, thisstudy will investigate the changing use of three parks placed throughurban revival planning projects: Maebashi Park, Shikishima Park,and Hirose Riverside park road.
By evaluating the features of parks planning and subsequent parkuse, this study considers present issues and future focal points inpark planning. This study has focused on three parks in Maebashi
City's urban revival planning projects, and considered the changing
use of these parks. It has found that all three parks began by opening
up scenery, and by clarifying park recreational functions they havebecome large-scale facilities for public gatherings.
However, while park recreational functions have increased, there
is also the possibility that the original intent for a connection with thescenery has been diluted.

Therefore, in future park preparation, even as partial
improvements, three considerations are significant in landscapedesign: (1) the origin of the natural scenery and land, (2) integrating
natural features and nature with design, and (3) subsequent use of thepark.
Keywords: Maebashi City, change of using, Urban Park
1. INTRODUCTION
This study looks at Maebashi City in Gunma Prefecture, one
example of a local city with a stock of many parks from post-war
urban revival planning projects, and aims to consider future parkdevelopments by evaluating park planning and use through anunderstanding of (1) the features of park arrangement in urban
revival planning projects, and (2) the changing use of planned parks.
First, this study will demonstrate the features of park arrangement
planning in urban revival planning projects. Next, this study will
investigate the changing use of three parks placed through urban
revival planning projects: Maebashi Park, Shikishima Park, andHirose Riverside Park Road. By evaluating the features of urban
park planning and subsequent park use, this study considers present
issues and future focal points in park planning.

敷島公園Park
Shikishima

Hirose Riverside Park Road
広瀬川公園道路

前橋公園
Maebashi Park

Higashi
Park
東公園

Figure.1 War-devastation urban revival planning in Maebashi City[1]*Maebashi City Office **Gunma National College of Technology ***Maebashi Institute of Technology

517

Picture.2 Maebashi Park[3]Picture.1 Maebashi Park[2]2. Features of Park Planning in Urban Revival Planning
Projects in Maebashi City
The Japanese park system began by official notice of the
Department of State in 1873. Planned preparations were made withthe promulgation of regulatory reforms in Tokyo in 1889. Throughthe promulgation of the former City Planning Law of 1919,
opportunities arose for securing planned parks even in local cities
through the use of "rezoning projects."
However, genuine park preparation began with the urban disasterrevival planning projects that followed the World War II, and theenactment of the City Park Law of 1956. The city planning ofMaebashi City progressed with the promulgation of the city
Planning Law of 1919, and in 1929 city planning zones were
determined. These brought opportunities for city planning projects
in Maebashi City. However, because approximately 80% of the city
was destroyed in the world War II, city planning projects could not
be carried out.
In 1946, Maebashi City was designated a war-devastated city, andthis was used as an opportunity to examine city plans from before thewar, and to move forward with road and park planning.
The area of rezoning projects for war-devastation urban revival
planning projects in Maebashi City was set in 1946 at 343.2 hectares,
and the cabinet approved a plan for major thoroughfares. In 1947,
Gunma Prefecture officially announced route 120.
This plan included three items regarding parks in the basic policyof city revival planning. First is the planning of parks, sports grounds,
and park roads. Second is the securing of over 10% of urban area as
open space. Third is the designation of agricultural land, mountainand forests surrounding the urban as open space. Park arrangement
was attempted under these three items.
However, citizens protested, demanding the prefecture "make it
Maebashi City's original planning. Therefore, the basic policy forpost-war urban revival planning was changed in 1949. The aim
became quick execution of projects.
Already in Maebashi City preparations had been made for
Maebashi Park along the left bank of the Tone River flowing
north-south through the city center, and for Shikishima Park.
However, due to the rising demand for open spaces and tree, park
plans through war-devastation urban revival planning projects were
proposed.
First, Maebashi Park was prepared for the west side of Maebashi
City, and Higashi Park was planned for the east side of the city. Also,a park road was planned for Hirose riverside flowing east-west
through the city, connecting the parks (figure 1).

3. Changing Use of Maebashi ParkThe establishment of Maebashi Park was decided by resolution of
the Maebashi city council in 1903 in commemoration of the
Russo-Japanese War of 1905.
It is recorded that this place was used for recreationalcherry-blossom viewing. Cherry-blossom viewing was a popular
outdoor recreation in Japan from the Edo period. Records of park use
show a design of cultivated cherry-blossom trees and garden-likeviews with additional trees. Monuments were erected, increasing
symbolism, and the park was used as a place for holding national
events.
Female factory workers and geisha used the park as a place toenjoy cherry-blossom viewing and strolls. With symbolism in its
scenery from the Meiji era, and use for strolls and events, everyday
use of Maebashi Park increased, and the convenience of the facility
improved with the arrangement of gazebos and flower beds.
Everyday use of the park increased, and a boat dock facing theTone River was built. Neighboring farms were bought, and the parkwas expanded.
Waterfalls and fountains (picture 1), and children's playing
equipment were built in the expanded areas. On the other hand,management problems such as public disturbances, crime, andsafety appeared in the park. Lighting was set up to prevent crime,and moral edification was conducted in response to vandalism offacilities. Records show that bicycle competitions and plays wereabundantly held. During public gatherings, rice cakes called"hagi-mochi" were sold in the park.
Accompanying the park's change from scenery viewing to moredaily use, park use became more integrated with daily life, andmodern culture such as food and the arts ripened.
In 1946 at the start of urban revival planning projects, Maebashi
Park became a normal park of approximately 18.6 hectares. In 1954,
to celebrate the 60th anniversary of the founding of Maebashi City,
the Maebashi Ground Fair was held.
In 1986 to celebrate the 100th anniversary of the founding ofMaebashi City, a plan for improving Maebashi Park was revealed
that would make the park a general park of 64.4 hectares, and even atpresent plans for refurbishing the park are being advanced (picture2)[6].

4. Changing Use of Shikishima Park
Shikishima Park lines the clear waters of the Tone River, rich inscenery symbolized by stretching pine forests, and blessed withviews of the Three Jōmō Mountains of Akagi, Haruna, and Myōgi.

518

Picture.3 Shikishima-Park[4]Picture.4 Shikishima-Park[3]In 1921, an outdoor school was created in the summer in
Maebashi City using pine forests, and it served as a place for
improving the constitution of children. In 1922, the Maebashi City
Council resolved to make the land a park site, and the park was madeby disposing of the riverbed and national forest.
Such was the origin of Shikishima Park, which opened an areathat included a riverbed blessed in natural scenery to the public as arecreational garden.
In 1925, a local youth labor service built a sports ground. The parkname of "Shikishima Park" was decided by accepting submissions
from Maebashi citizens. In 1924, citizens donated Yoshino
Cherry-blossom, improving park scenery. In 1929, Dr. Honda
Seiroku and Dr. Inoshita Kiyoshi proposed a lake with islands as
seen in the Chinese Imperial Garden, a stone-made rest area
imitating a French garden, and toilets (Picture 3).
In a place blessed with a natural environment favorable enough tohouse an outdoor school, athletic facilities were enriched.
In 1930, a baseball field was built. In 1920, U.S. and Japanese
professional baseball teams met for a friendly game that including
Sawamura Eiji and homerun-king Babe Ruth. In 1938, municipal
rental boats were opened for people to enjoy the scenery.
In 1950, the grounds were expanded for the construction of a

Gunma Prefectural General Sports Ground. In 1953, a volunteer
park beautification club was organized of townspeople to clean thepark, and townspeople conducted cleaning activities. In 1956, Meiji
Shine conveyed 500 irises, and an iris garden was created. In 1971,
with the creation of a rose garden, the institutionalization of the park
as a tourist resource progressed. In 1983, a national sports festival
was held at the sports grounds of Shikishima Park (Akagi Kokutai).
In 1985, with the addition of a drainage channel, a water playground,and a large wooden playground in the connection point between therecreational walking zone and the sports facility zone, the park wasgiven general functions that various generations could use (Picture4). In 1989, Shikishima Park was chosen by the Parks & Open Space
Association of Japan as one Japan's top 100 parks.[7]5. Changing Use of Hirose Riverside Park road
The Hirose River was a part of life during the Edo period as atransportation route for boats, and in the early Meiji period powering
water wheels. The city center side was mainly used as commercial
land, and the new city center side was mostly used as industrial land(silk factories). On holidays, many female factory workers wouldcross the Hitone Bridge to enjoy the city center (figure 2).

Figure.2 Shinkeizu in Maebashi city[5](1909)
519

Picture 5 Hirose Riverside Park Road[2]Picture 6 Hirose Riverside Park Road[3]Hagiwara Sakutarō, who influenced modern poetry, announced
his poem "Kyōdo Bōkyōshi" in 1925, which includes reference toHirose River, and he is said to have liked this scenery. In 1930,
preparations began for Hirose River park road, 3.6 meters wide and283 meters long. In the early Shōwa period, Hirose riverside wasprepared as a street.
Following the end of the World War II, in 1948 Maebashi Park
(general park) was planned to the west of the city center, Higashi
Park (normal park) was planned to the east of the city center, and"Hirose Riverside Park road" was planned as a park road connecting
the two parks in light of city disaster prevention.
The land for "Hirose Riverside Park Road" was secured throughrezoning projects for the purpose of urban disaster prevention in
Maebashi City's central areas. Weeping willows were planted ateven intervals on both banks, and a dirt promenade along the rightbank was built. A concrete bench was built along the promenade,and a green tract of land was built in a concrete tree-planting box.
Through this preparation, the weeping willows planted alongside
both banks of the river became familiar to citizens as the scenery
representing Maebashi City.
In 1970, a monument inscribed with a poem entitled "Hirose
River" was built. This was followed by many other hometown poetic
monuments erected in the promenade. In 1974, Maebashi City
enacted the "Maebashi City Water and Green Community
Development Ordinance," and for five years from 1974 to 1979
reconstructed Hirose Riverside Park road as a cultured andhigh-quality space, with the theme of making a symbolic road ofwater and tract.
Inside, a stone walking path was added, a viewing lake andmonument were built, and preparations focusing on recreation andscenery-viewing progressed.
After this, cultural hues were deepened in the areas surrounding
Hirose Riverside Park Road, and in 1993 the Maebashi Museum of
Literature was built along the Hirose Riverside Park road, with materials
from poets on display and for research including the work of Hagiwara
Sakutarō. Every year since 1999, from autumn to winter both sides of
the riverside promenade are lit with lights by volunteers, making anelegant scene[8].

6. CONCLUSION
This study has focused on three parks in Maebashi City's urban
revival planning projects, and considered the changing use of theseparks. It has found that all three parks began by opening up scenery,and by clarifying park recreational functions they have becomelarge-scale facilities for public gatherings.
However, while park recreational functions have increased, there
is also the possibility that the original intent for a connection with thescenery has been diluted.
Therefore, in future park preparation, even as partial improvements,
three considerations are significant in landscape design: (1) the origin of
the natural scenery and land, (2) integrating natural features and nature
with design, and (3) subsequent use of the park.

7. REFERENCES[1] Maebashi City (1964): War-Devastation Urban Revival Panning in
Maebashi City.
[2] Hirata, K. (1990): Taisho Period from age of the Showa era.
[3] Maebashi convention :http://www.maebashi-cvb.com/tourism/view/i-kanko/03k-maebashi.htm[4] Taro S. (2004): The history of Iwagami.
[5] Gunma Shinbunsya (1911): Shinkeizu of Maebashi-city.
[6] Tsukada, S., Yuzawa. A., (2004): A Study of the Evaluation Structure andAttractive Factors of Large-Scale Park caught from User's Consciousness-A case Study in Maebashi City-, City planning review, Special issue,
Papers on City Planning(39), pp.193-198.
[7] Tsukada, S., Morita, T. and Yuzawa, A.(2009): A study on evaluation of
Shikishima park as change of the spatial characteristic and the original
design, journal of the Japanese Institute of Landscape Architecture72(5),pp.849-854.
[8] Tsukada, S., Morita, T. and Yuzawa, A.(2011): The Evaluation of theplanning Designs in Hirose Riverside Promenades, journal of the JapaneseInstitute of Landscape Architecture 74(6), pp.18-21.

520

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

A Study on Evaluation of Quality of Life in Consideration of
Water/Green Environment
Tetsuo Morita*, Yoshihito Kogure*, Hiroshi Sugita**,
Tsuyoshi Baba**, Shin-ya Tsukada***, and Naoki Miyazato**Gunma National College of Technology, **The Institute of Behavioral Sciences, ***Maebashi City Office

ABSTRACT: Recently, the interest about water/green environment
rises. But, it does not become clear enough, that water/green
environment influences Quality of Life. The purpose of this study isthe analyzing Quality of Life quantitatively with the data of thequestionnaire survey, from the viewpoint of water/green
environment. The subject area was Maebashi City, Gunma
Prefecture. We used “factor analysis” and “covariance structure
analysis”. Result of factor analysis we obtain five factor of the
Quality of Life: Safety, Convenience, Environment, Housing
conditions, Comfort. In addition we clarify evaluation structure of
Quality of Life in consideration of water/green environment such as
the river or the park by using covariance structure analysis. As theresult, we made clear that Quality of Life is different depending on
individual attribute and district characteristic, and that water/green
environment affected Quality of Life. And we show measure effects.

Keywords: water/green environment, Quality of Life,evaluation, covariance structure analysis
1. Research Background and Purpose
In these modern times, with advancing industrialization andurbanization, the scenery of the water and green environmenthas been transformed. People in these present times, have aheightened desire, a craving for spaces where they can live atease. Their interest for the Quality of Life with as muchnatural contact with those familiar green spaces, waterfront
and the beautiful townscape has increased. In this research,based on the results of a questionnaire survey regarding the
Quality of Life of the residential areas of Maebashi city,
special attention is paid to the living environment of theresidential areas that incorporate water/green and the like.
Our purpose is a making structural model, for the evaluationof Quality of the Life, as seen from the inhabitant’s point ofview.
2. Research Position
In the research that pays attention to the Quality of Life,
Kaji’s research is enumerated [1]. They clarify structural
analysis of the consciousness of the inhabitants. Moreover,they showed the relation between living environment andevaluation of inhabitants to living environment, and offer an
evaluation method of their living environment and the
Quality of life. After this research, various evaluation
methods and models have been proposed by Morimoto [2],Yoshida [3], Doi [4], and Morita [5]. In addition, a Quality ofLife research which pays attention to water environment,
Taniguchi is also included [6]. They propose that in order toimprove the quality of life, an improvement in the water
environment and a better evaluation of this model need to be
made. Not just as research, a plan which designates
improvements to the charm of the water spaces in Tokyo is set
out as a goal [7]. Such as this plan, many urban development
plans that pay attention to the water environment are going on
all over Japan.
The main focus of this research is on the Quality of Life
especially to the living environments with water and green.
The Quality of Life of the inhabitants is determined by city
activity and living environment. City activity consists oftraffic, land utilization and economic activity etc, living
environment consists of the proximity to rivers, forested land,
farmland and the growth plants and other living things. These
two are thought to exert mutual influence on each other. In
addition, it is clarified that the Quality of Life is different
depending on individual attributes and district characteristics
by existing researches. In this research, individual attributesand district characteristics are considered. As we a model
which evaluate all the Quality of Life and which pays
attention to a water/green environment has been established.
This research aims to provide a useful model through which
all aspects of quality of life can be measured quantitatively
especially focusing on the water/green environment.
3. Questionnaire Survey of Quality of Life
3.1 Research Area
The area in which the Quality of Life questionnaire survey
was conducted is designated as Maebashi city, Gunma
prefecture. Maebashi city has three “class A rivers”: Tone
river, Hirose river and Momonoki river. In addition, thehydrophilic park which begins in the Shikishima park locatedin Maebashi city. And, Maebashi city advertises itself as:“The city of water, green and poetry”. Maebashi city
promotes businesses such as promenades along the rivers aswell as other services and the reforestation an improvement of
the cycling road following the plan which is called “Thegeneral plan for the Greening of the Maebashi city” [8].
Specially, we designated research area as Maebashi city of
the left bank of Tone river that is affected by water
environment. Research area is shown in Fig.1 on the next
page.

521

Distribution:Mid-November, 2008
Collection:December 14, 2008 (Mailing deadline)
Research Area Maebashi City of the left bank of the Tone River
Population:99,000 Household number:42,000
Systematic random sampling to household
Sampling
Sampling rate:9.5%
Object persons Members of 4,000 households
Survey method Distribution:Posting
Collection:Mailing
Survey items 1) Parsonal attribute, Household attribute, House style2) Five stage evaluations of Quality of Life
(1.Very good, 2.Good, 3.Average, 4.Poor, 5.Very poor)
Sample number Household:1,293 (Recovery rate:32.3%)
Pesons:2,118 Valid vote:1,646
Date

Table 1: Summary of the questionnaire survey

Fig.1: Research Area
3.2 Summary of Questionnaire Survey
Summary of questionnaire survey is shown in Table 1, anditems of evaluation are related the Quality of Life (Subjective
evaluation) is shown in Table 2. The survey was distributed to
4,000 households inside the area of our research, we received
2,118 responses. When those responses where appraised andreviewed as being valid and complete, the effective results
were reduced to 1,646. Items of Subjective evaluation (Table2) refer to appraisal items of Quality of Life research by [3]and [5], it added the item regarding water and green
environment (A12, A13).

Variable
Variable name (Survey items of QoL)
A1 Safety against the flood damage
A2 Safety against the earthquake and fire
A3 Crime prevention
A4 Safety against the traffic accident
A5 Sanitary condition
A6 Noise and vibration
A7 Accessibility to the post office and bank
A8 Convenience for the commuting
A9 Accessibility to the hospital and welfare facility
A10 Convenience for the shopping
A11 Convenience for the public transportation
A12 Prodigality of the green environment
A13 Prodigality of the water environment
A14 Convenience for the sport and recreation
A15 Goodness of the sunshine and ventilation
A16 Extent of the house and garden
A17 Easiness of the walking in the district
A18 Goodness of the townscape
A19 Usability of a car in the district
A20 Usability of a bicycle in the district
A21 Comprehensive evaluation

Table 2: List of Subjective evaluation

4. Factor Analysis

Factor1

Factor analysis was applied to items of Subjective evaluation,
the component of Quality of Life was extracted. With thisanalysis, because A17 and A20 is correlatively high which isshown in Table 2, it excluded the item of A20. Also A21 was
excluded of analysis. Concerning items of Subjective
evaluation, the typical five latent variables where sum ofsquares exceeds 1.0 were extracted. The Table 3 is somethingwhich rearranged the factor loading after the varimax rotation.
We defined factor collected those where the respective factor
load quantity is high, “Safety” and “Convenience”,“Environment” and “Housing conditions”, “Comfort” . A12 ,A13 which are thought that the relation to green and water
environment is high during Subjective evaluations,“Environment” in the Quality of Life which is defined, it has
the respective factor loading 0.821 and 0.737. It understoodthat the element regarding water and green is related to the
Quality of Life largely.
522

Variable
A1
A2
A3
A4
A5
A6
A7
A8
A9
A10
A11
A12
A13
A14
A15
A16
A17
A18
A19
Sum of squares
Contribution ratio
Cumulative
contribution ratio

SafetyFactor2

Factor3

Convenience Environment

Factor4
Housing
Conditions

Factor5
Comfort

0.765
0.750
0.602
0.598
0.441
0.373
0.109
0.049
0.145
0.082
0.157
0.130
0.252
0.209
0.194
0.233
0.324
0.291
0.294
2.761
14.53%

0.139
0.067
0.185
0.103
0.277
0.081
0.731
0.723
0.698
0.577
0.541
0.064
0.058
0.250
0.100
0.079
0.303
0.169
0.358
2.656
13.98%

0.036
0.183
0.156
0.249
0.285
0.323
0.090
0.029
0.121
0.053
0.069
0.821
0.737
0.504
0.174
0.251
0.169
0.270
0.164
2.029
10.68%

0.136
0.211
0.081
0.104
0.179
0.290
0.042
0.063
0.050
0.126
-0.013
0.103
0.276
0.139
0.723
0.688
0.158
0.299
0.239
1.494
7.87%

0.060
0.146
0.132
0.236
0.161
0.110
-0.0220.155-0.011
0.159
0.232
0.093
0.095
0.174
0.079
0.224
0.623
0.504
0.390
1.138
5.99%

14.53%

28.51%

39.19%

47.05%

53.04%

Table 3: Result of Factor analysis

5. Presumption Evaluation Model of Quality of Life
5.1 Covariance Structure Analysis
In main research of the Covariance structural model, when
Quality of Life is appraised, we use Yoshida’s model [3].
This model can analyze the individual attribute, district
characteristic, Subjective evaluation, and Quality of Life.
Covariance structure analysis can display the complicatedstatistical model graphical causal relation between variables
the arrow (path) with by the path figure which is displayed. It
is possible because with the structural equation model to form
the model which is based on the hypothesis of the analyst incomparison with former multivariate analysis, it is flexible tointerpret data which is given.
Individual attribute and district characteristic were designatedas Objective variable, those Latent variable (Quality of Life)
exists also constructed the Covariance structural model
between Subjective evaluations. Causal relation between
Objective variable and Latent variable is called structural
equation, this is suitable to multiple regression analysis. In
addition, causal relation of Latent variable and subjective
appraisal value is called measurement equation, this issuitable to factor analysis.
5.2 Presumption Covariance Structure ModelContinuously, considering five Quality of Life, it set the caus
al relation of individual attribute and district characteristic, also presumed the Covariance structural model. We obtain dat
a of district characteristic on the map. Goodness of fit of themodel became GFI=0.80. Sufficient goodness of fit you cannot say, but sign condition of path coefficient, the appropriaten
ess of interpretation of variable, it illuminated in purpose of t
his research which pays attention to water/green environment
the figure adopted the model of Fig.2 on next page. The Tabl
e 4 is summary of Objective variables, and the Table 5 is pat
h coefficient of the variable which is used for the model is sh
own in Fig.2. Table 4 and Table 5 are on the next page.
“Safety”, the flood damage, earthquake, fire, crime preventio
n and traffic accident, is the Latent variable which is related to the safety regarding hygiene. It is found that from the fact t
hat path coefficient of the dummy of 65 years old or more ofobjective variable has shown negative value, as for the seniorcitizen degree of safety decreases. “Convenience” is the Lat
ent variable which is related to the convenience of shopping
and public traffic. From the fact that path coefficient of the ages 65 and older dummy and the employee dummy of Object
ive variable has shown negative value, the senior citizen andthe employee is a tendency where convenience decreases. On
the one hand, path coefficient of the student dummy has sho
wn correct value, as for the student, it is found that it is the tendency which appraises the convenience of research area high.
“Environment” is the Latent variable which is related to water/green environment which has paid attention in this researchand sport recreation. As Objective variable is distance to thecity park, path coefficient it reaches negative value, the city p
ark becomes far, as for appraisal of “Environment”. It is found that it decreases.
Path coefficient of the employee dummy and the apartment
dummy where “Housing conditions” is the Latent variablewhich is related to sunshine, through wind sequence and theextent of the house and garden, is Objective variable has
shown correct value. “Comfort” is the Latent variable whichis related to the easiness of walking, easiness of using of theautomobile. Objective variable becomes the variable whose
distance to the waterfront is significant, has shown the factthat extent and the comfort where distance to the waterfront
becomes far decrease. Like above, to improve the Covariancestructural model is also conversely.
In Objective variable, the fact that it is related to water/green
environment is distance to the city park and distance to thewaterfront. Waterfront such as Tone river, Hirose river andMomonoki river, and city parks has been distributed toresearch area, it became the model which can explain the
Quality of Life of residence area with the approach
characteristic to water/green environment. In addition, also it
was the model which can explain the Quality of Life withindividual attribute, concerning the change of Quality of Lifedue to the change of future population became the model
which can be appraised.
At the time of Covariance structure analysis of this research,
the fact that Amos of the SPSS co. is used.
6. Evaluation Measures
By using Covariance structural model, evaluate water green
environmental measure. The next two measures, “waterfront
development” and “City parks development” were set. Fig.3
shows positions about new waterfront and city parks on thenext page but one.
6.1 Waterfront Development
Such as Tone river, Hirose river and Momonoki river which
are existence as waterfront in research area extension is
34.5km. In addition, there is a irrigation canal 29.1km insideresearch area, presently, it is the space which the resident
can’t get close, part has become the culvert. These like theHirose river and the Momonoki river you service as thewaterfront inside the city, it makes the space where theresident can become familiar in the waterfront. Because ofthis, rivers extension inside research area reaches
approximately 2 times, it is supposed that “Comfort”
improves due to the fact that distance to the waterfront
becomes short from each area.
6.2 City parks Development
67 city parks are maintained inside service research area, thecity park exists in inner 47 areas of 84 areas inside researcharea. Among 37 areas where the city park is not serviced, 6
areas of urbanization control area are excluded, the city park
(the block park) to set one each in 31 areas of area designatedfor urbanization. Because of this, the number of city insideresearch area parks becomes with 98 and approximately 1.5
times from 67, “peripheral environment” improving is
supposed due to the fact that distance to the city park becomes
small from each area.

523

Fig.2: Estimated Covariance structure model
Division
Z1
Personal
Z2
attribute
Z3
Z4
DistrictZ5
characteristic Z6

Definition
Variable name
Ages 65 and older dummy Sample aged 65 and over = 1, the others = 0
Student dummy
Sample is a student = 1, the others = 0
Employee dummy
Sample is an employee = 1, the others = 0
Apartment dummy
Sample lives in the apartment = 1, the others = 0
Distance to city park
Square of distance to city park (square kilometer)
Distance to waterside
Square of distance to waterside (square kilometer)

Table 4: Objective variables
Objective
variables
Z1

Z1
Z2
Z3

Z5

Z3
Z4
Z6

SubjectivePathevaluation
coefficient
0.754(18.526)
A1
0.802(18.927)
A2
0.643(17.231)
A3
Safety0.677(17.670)
A4
0.557(15.935)
A5
0.492( ∞ )A6-0.070(2.536)
0.729(19.973)
A7
0.058(2.137)
0.725(19.992)
A8-0.057(2.089) Convenience 0.696(19.506)
A9
0.609(18.030)
A10
0.562( ∞ )A11-0.076(2.793)
0.863(20.881)
A12
Environment 0.792(21.611)
A13
0.575( ∞ )A140.088(4.291)
Housing
0.516(3.7350)
A15
0.116(5.626) conditions
1.172( ∞ )A16-0.110(3.913)
0.838(18.232)
A17
Comfort0.656(19.455)
A18
0.605( ∞ )A19
The value in round brackets is “t value”.
Pathcoefficient-0.185(6.656)

Latentvariables

Table 5: Path coefficient524

Fig.3: Waterfront and City parks Development

Fig.4: Result of Waterfront development

Fig.5: Result of City parks development525

6.3 Evaluation of Water/green environmental Measures
The result of measuring of water/green environment, measure
improves each one “Comfort” “Environment”. Increase ofevaluations of “Comfort” and “Environment” after thewaterfront and city parks development shows Fig.4 and Fig.5
on the previous page. In order to grasp regional the effect tothe whole research area, population of classified by appraisal
value ranking of Quality of Life was rearranged. The Fig.6
shows change of “Comfort” due to the service of thewaterfront. The Fig.7 shows change of “Environment” due tothe service of the city parks. Sooner or later the population
which enjoys the improvement effect of quality of life withthe measure which relates to water/green, has increased.

7. Future Topic
Future topic is improvement the evaluation model of Qualityof Life. In this research, water quality content of the rivers
and waterfront as water environment is not included. In
existing research, a model involves the water cycle, matter
cycle model which can estimate the change of water quality
and water content. In the future, we would like to examine theappraisal Quality of Life with water quality and water content.
In addition, in order for appraisal of measure other thanwater/green environment to become possible, we would liketo keep adding various district characteristic.
8. REFERENCES
[1]

[2]

[3]

[4]

[5]

Fig.6: Change of “Comfort” due to measuring
[6]

[7]
[8]

Fig.7: Change of “Environment” due to measuring

6. Research Conclusion
Summary of Objective variable of this research (individual
attribute and district characteristic), Latent variable,
Subjective evaluations which is due to questionnaire survey
uniformly, also it was possible to construct the covariancestructural model. It was constructed individual attribute andthe district characteristic regarding water/green form the
Quality of Life. When appraisal of waterfront and city parks
service measured which relates to water green environment.
We verified that “Comfort” and “Environment” improves. Asfrom above, the purpose in this research cleared that theappraisal structure of Quality of Life which includes
water/green environment, it is thought that it was possible to
do measure appraisal.
526

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Mamoru, T. Hiroaki, H. et al. “EVALUATION OF
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First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Changes and Issues in Green Space Planning in the Tokyo Metropolitan
Area: Focusing on the "Capital Region Plan"
Tetsuo Morita*, Yoshihide Nakagawa**, Akinori Morimoto***,Masateru Maruyama*** and Yoshimi Hosokawa**Gunma National College of Technology, **Waseda University, ***Utsunomiya University,****Mizuho Research Institute

ABSTRACT: Large-scale green space planning in theTokyo Metropolitan Area was formulated in 1939 with the"Tokyo Green Space Plan," and a greenbelt was proposed,
however the plan was cancelled due to the end of the war. In
1958 the "First Capital Region Plan" was formulated, and agreenbelt was again proposed, however the second and laterplans became plans that approved of urbanization.
This paper focuses on the Capital Region Plan, and aims to(1) organize changes in green space planning in the Tokyo
Metropolitan Area, (2) analyze the circumstances ofurbanization of the area designated a greenbelt in the First
Capital Region Plan, and (3) extract issues regarding green
space planning in the Tokyo Metropolitan Area
Keywords: Tokyo Metropolitan Area, Capital Region Plan,
National Plan, greenbelt

1. Changes in Green Space Planning in the Tokyo
Metropolitan Area
1.1 Tokyo Green Space Plan
In 1939 the "Tokyo Green Space Plan" was formulated.
Although this planned area had no legal basis, it included a 50
kilometer region, 962,059 hectares of production green
spaces and scenic districts centered around on Tokyo. A
greenbelt was planned for Tokyo's outer region, following
rivers into urban areas[1].
The base section of the greenbelt was set in the city plan asa planned city green space, and preparations were madethrough site purchases. In 1940, with the war escalating,Tokyo industrialized six green spaces for air defense, and by1945 had realized the plan by deciding on 22 green spaces forthe city plan. The air defense designation prohibited theconstruction of buildings other than those affiliated with
agriculture, forestry, livestock, or park sports grounds.
However, after losing the legal backing of air defense withthe end of the war, these large green spaces became a target
for opening agricultural space, and 62% of the 746 hectares ofpurchased sites were transferred to private landowners.
1.2 First Capital Region Plan
Due to the rapid concentration of population in Tokyo
during the period of high economic growth, chaotic urban

expansion became a problem, and the National Capital
Region Improvement Act was enacted (1956). Based on thisAct, the first through fifth National Capital Region
Improvement Plans (abbreviation: “Capital Region Plan”)
were formulated, aiming to facilitate the construction andordered development of a metropolitan area suitable to be thecenter of Japanese politics, economics, and culture. The First
Capital Region Plan of 1958 referenced the Greater London
Plan, creating Suburban Area (Greenbelt), and aimed to"control the chaotic and swollen development of Built-up
Area, and facilitate healthy development by creating an outergreenbelt." The greenbelt was planned on the outer side of
Built-up Area, with a width of 10 kilometers, and satellite
cities were planned outside the greenbelt (Figure 1).
However, the understanding of municipalities and landowners of the areas designated as Suburban Area was notgained, and improvement was made difficult by opposition
activities of residents. Furthermore, the land use planningsystem regarding green spaces was left incomplete. In
addition, achievement of the plan became difficult with the
chaotic development of Suburban Area in the 1960s due tothe intense concentration of population and the rapid
expansion of urban areas, making change unavoidable..
1.3 Second Capital Region Plan
The Second Capital Region Plan was formulated in 1968
(Figure 2). The Suburban Area (Greenbelt) of the First
Capital Region Plan was changed to Suburban Development
Area "in order to prevent chaotic urbanization throughplanned maintenance of urban areas and preservation of green
spaces," and the concept of a greenbelt faded. Most of theurban areas in the 50 kilometer zone designated as Suburban
Area in the First Capital Region Plan became Suburban
Development Area meant for commuting to Tokyo.
The greenbelt concept partially remained. Among thegreen space in the Suburban Development Area, areas indanger of chaotic urbanization were designated Suburban
Green Space Preservation Area (18 zones, 15,693 hectares).
Furthermore, 9 districts, 758 hectares of Suburban GreenSpace Preservation Area that were considered important
districts especially in need of preservation were designated
Suburban Green Space Special Preservation Area, making
permission of the governor necessary before conducting
certain development activities.

527

Suburban Area
(Greenbelt)

Built-up Area

Figure 1: First Capital Region Plan[2]Suburban Development
Area

Built-up Area

Figure 2: Second Capital Region Plan[2]5282. Changing Land Use of Suburban Area over Time
This section takes up Kawaguchi City in Saitama
Prefecture, planned in the First Capital Region Plan as
Suburban Area (Greenbelt), and analyzes its subsequent land
use over time. Kawaguchi City contains both a designatedBuilt-up Area and a designated Suburban Area[4] (figure 3).

Residential zone
Category I exclusively residential zone
Category II exclusively residential zone

Built-up Area

Greenbelt
Kawaguchi City

Commercial zone
Neighborhood commercial zone

Urbanization
Promotion
Area

Quasi-industrial zone
Industrial zone
Exclusively industrial zone

Figure 3: Kawaguchi City and Greenbelt

Figure 5: Percentage of Various Land Use Plan of Built-up
Area (Kawaguchi City)

2.1 Changes in Land Use Planning
In 1963, five years after the formulation of the First Capital
Region Plan, approximately 20% of Kawaguchi's Suburban
Area was designated an urbanization area, however in 1970,
with the increased demand for housing due to rapid
population growth, approximately 85% had become an
urbanization area. As a result, Kawaguchi City had nearly as
much designated urbanization areas as existing urbanization
areas (figures 4 & 5). Compared to Built-up Area, the ratio ofland use of Suburban Area for residential uses had greatly
increased.

Residential zone
Category I exclusively residential zone
Category II exclusively residential zone
Commercial zone
Neighborhood commercial zone

Urbanization
Promotion
Area

Quasi-industrial zone
Industrial zone
Exclusively industrial zone

Figure 4: Percentage of Various Land Use Plan of the
Suburban Area (Kawaguchi City)

This indicates that Kawaguchi City, after the First Capital
Region Plan (1958) was revised by the Second Capital
Region Plan (1968), rapidly transitioned towards
urbanization areas, and its character as a Tokyo bed town
strengthened.
Viewed from a map (figure 6), in 1963, following theformulation of the First Capital Region Plan (1958),
approximately 80% of the Suburban Area was designated anurbanization control area.
In 1968, when the Second Capital Region Plan wasformulated, residential demand increased, and the designationof urbanization control areas for residential use can be seen.
In 1970, in the period between the formulation of the Second
and Third Capital Region Plans, approximately 70% of thearea that had been designated as Suburban Area in the First
Capital Region Plan had become designated for residentialuse. In 1985, when the Fourth Capital Region Plan wasformulated, 85% of the Suburban Area of the first plan had
been designated for use, leading to urbanization, and at thesame time an intention to halt the increasing population in
Suburban Area can be seen.
2.2 Changes in the Record of Land Use over Time
Looking at the change in the record of land use over time
(figure 7), in Built-up Area designated in the First Capital
Region Plan, there was an urban land use structure, withbuilding sites occupying approximately 80%. Looking at the
Suburban Area of the first plan, after the period of the Second
Capital Region Plan in 1976, building sites increased
annually, and the ratio of fields and rice paddies decreased.
This can be seen as due to Kawaguchi City assuming the roleof Tokyo's bed town, taking on the increase in housing in its
Suburban Area.
Population continued to increase in
Suburban Area, and green spaces (forests, fields, rice529

1963 (pop: 199,000)

Residential zone

グCommercial zone
Quasi-industrial zone
用途地域
住Industrial zone
商Exclusively residential zone
準Exclusively industrial zone

用Greenbelt

Built-up Area05

1 0 km1968 (pop: 275,000)

Residential zone

グCommercial zone
Quasi-industrial zone
用途地域
住Industrial zone
商Exclusively residential zone
準Exclusively industrial zone

用

0

5

10 k m

用

1970 (pop: 297,000)

Residential zone

グCommercial zone
Quasi-industrial zone
用途地域
住Industrial zone
商Exclusively residential zone
準Exclusively industrial zone

0

5

10 km

Category I exclusively residential zone
Category II exclusively residential zone

1973 (pop: 321,000)

用 途 Residential zone

Neighborhood commercial zone
Commercial zone
Quasi-industrial zone
Industrial zone
Exclusively industrial zone

0

5

10 km

Figure 6: Changes in Land Use Planning (Kawaguchi City)
530

paddies) in 1996 were approximately 30% of the overall
Suburban Area and decreasing annually.

Forests

Kawaguchi City, designated as Suburban Area, thepopulation was 144,000 when the First Capital Region Plan
was formulated, and 343,000 in 1976 at the formation of theThird Area Plan, a 2.4 times increase. The policy ofpreserving green spaces can be seen to have been unable toresist the trend for expansion of urban areas due to theconcentration of population in a short time. The area divisionsystem with regulatory power was introduced, however
urbanization control areas were not arranged on a broad scale,and green spaces were reduced in each municipality.
(b) Political Understanding and Popular Support for theGreenbelt Concept
The period when Suburban Areas were planned coincides
with Japan's high economic growth, which produced ahousing shortage due to the concentration of population inlarge cities, and political consideration for satisfying housing
demand took precedent over the greenbelt policy forcontrolling the expansion of urban areas. Furthermore, withthe social background of prioritizing the economy, there wasno understanding among municipalities and citizens, andregulation was difficult. Even among experts, arrangement
and concern for green spaces was not established. Because
there was no political or popular understanding, there was noagreement regarding the Suburban area, and the plan became
weak with no regulatory or legal effect.
(c) Substance of the Greenbelt Plan
The Suburban Area was 10 kilometers from the city center.
On the other hand, the greenbelt of the Greater London Plan
was 20 kilometers from the city center. Viewed from thiscomparison, the Suburban Area created in the Capital Region
Plan can be called too close to the city center. As a result only
partial green spaces remained. This can be understood to be
due to the fact that, in addition to not foreseeing the
population increase, the Greater London Plan and the Town
and Country Planning Law were not sufficiently studied, andthe experience of England was not made use of.

Fields3.2

Suburban Area
100%
90%
80%
70%
60%
50%
40%
30%
20%
10%
0%
1976

1987

1991

1997

Built-up Area
100%
80%
60%
40%
20%
0%
1976

1987

1991

1997

Water
OthersTransportation facilities
Building sites
Wastelands

Rice paddies

Figure 7: Changes in the Record of Land Use (Kawaguchi
City)
3. Issues in Green Space Planning in the Tokyo
Metropolitan Area
3.1 Verifying the Greenbelt Concept
The concept of a greenbelt in the First Capital Region Plan
was an admirable plan meant to sketch an ideal city, however
in reality this had little practical effect, and it remained asregional green spaces. This section will investigate thereasons why the greenbelt concept was not fully realized,
covering the formative concepts of the plan, land use
planning of the Suburban Area of Kawaguchi City, andchanging land use.
(a) Expansion of Urban Areas Due to Population
Concentration during the Greenbelt Concept Period
Following the war, the concentration of population fromrural areas to Tokyo can be called rapid and extraordinary. In

Issues in Green Space Planning in the Tokyo
Metropolitan Area
In 2005, the population of Japan began to decline, andfurther population decline is assured. The greenbelt conceptof the First Capital Region Plan could not resist the pressureof population concentration during the period of higheconomic growth, however its significance today is worth asecond look. In order to realize the creation of compact cities
with a small environmental impact, city planning withappropriate arrangement of green spaces should beconsidered.
During the period of the First Capital Region Plan,
residential policy took precedent over green space policy,
however today the significance of green spaces is being
reconsidered. The role of green spaces in creating scenery,
food production, and disaster prevention measures isincreasing. The administration, citizens, and experts sharethe value of green spaces, and there is a need to study theproper structure of cities and green space planning.
In addition to the Capital Region Plan, in 2009 the "Regional
Plan of Metropolitan Area" was formulated, and a network
connecting existing waterside space and green space has been531

proposed. This plan does not indicate a concrete method forpreserving green space. What is needed is the creation of agreen space plan in the Tokyo Metropolitan Area that is basedon a study of the experiences of the greenbelt plan in the
Greater London Plan, Japan's Capital Region Plan, and green
space planning in various countries, as well as anunderstanding of present land use in Japan, and future
population and economic trends.
4 REFERENCES
[1]
[2]
[3]
[4]

[5]

[6]

Committee of Metropolitan Area (1959), Metropolitan Area Research,
Tokyo. (in Japanese)
Suzuki, S. (1993) Current of City Planning -Tokyo, London, Paris and
New York-, Sankai-do, Tokyo. (in Japanese)
Sato, A. (1977) History of the Development of Japanese Park GreenSpaces (upper) (lower), Chikuma-syobo, Tokyo. (in Japanese)
Maruyama, M. Nakagawa, Y. (2004) A study on the changes of green
space policies of National Capital Region Improvement Plan,
Proceedings of annual conference of the Japan Society of Civil
Engineers, Vol59(4), pp334-335.
Takeuchi, T., Ishikawa, M. (2009) A study on the effect of urban
development projects on the open space environment in areas once
designated as green districts, LRJ, 72(5), pp.705-708.
Takeuchi, T., Ishikawa, M. (2009) A study on the green space policies
on the planned areas for Green Belt in the National Capital Region
during the period of urban expansion -a study in the North Tama Area-,
Journal of the City Planning Institute of Japan, No.44-3, pp.877-882.

532

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Application of Traffic Simulation in Road Policy Assessment Accompanied by
Railway Elevation
Takuya MAEKAWA1) ・Hiroshi INOUE2)
1)

2)

Graduate School of Human Science and Environment, University of Hyogo
maekawa@maruokeikaku.co.jp

Professor, School of Human Science and Environment, University Hyogo, 1-1-12,
Shinzaikehoncho, Himeji,Japan
inouye@shse.u-hyogo.ac.jp
behavior of traffic flow on the roads. 2) Traffic demand isalso given to the model as a dynamically changing origin
and destination traffic volume matrix. The output of themodel is ever changing traffic flow condition shown
visually. Also quantitative data on traffic volume, length ofcongestion travel speed and exhaust emission are provided
as output of the model. These output of the simulator, visualgraphics output and quantitative information, make it easy
to explain and understand the effect of transportation and/or
land development projects.



ABSTRACT: In road policy assessment, it is very important topresent the influence and effect of the project simply, and to holdinformation in common with relatives. Thus, traffic simulation
plays a part in road policy assessment. In this paper, we present
how to deal with traffic simulation and introduce an example of
the application, in Road Policy Assessment accompanied Railway
Elevation. Lastly, we show possibility of traffic simulation inRoad Policy Assessment.

Key Words: traffic simulation, road, road traffic, traffic
flow, road policy, road policy assessment, railway elevation
1. INTRODUCTION
Traffic simulation is a tool to represent traffic flow on theroad network by use of computer, and to estimate theinfluence of new road or land development on the traffic
flow beforehand. The result of the simulation is represented
as visual information in the series.
In recent years, public involvement is introduced totransportation policy on the view point of promoting
democracy and transparency. As relatives contained local
residents are likely to get mutual understanding, if
information such as traffic congestion and/or the length ofcongestion is shown as ever changing visual scene. Whence,
the meaning of traffic simulation is recognized as a tool ofmaking mutual consent for promoting transportation policy.
In this study, we try to verity the usefulness of traffic
simulation for promoting road policy in urban area.

Fig.1 Input and Output of the Traffic simulation “VISION”
2.2 Traffic Flow Model of the Simulator “VISION”
In the traffic simulator “VISION”, we use car-following
theory as a traffic flow model. 3) The acceleration of avehicle is determined by a function of the gap distance withthe preceding car and the difference between desired speedof the car concerned and the actual speed of the precedingcar.
The function is given by

2. OUTLINE OF TRAFFIC SIMULATOR “VISION”
2.1 Model Structure
Traffic simulator “VISION” consists of traffic flow modeland route choice model.1) Traffic flow model determine
behavior of each vehicle such as speed and vehicular gap on
the condition of those at previous time.
By the repetition of the procedure, the condition of traffic
flow on the road network is refreshed periodically (periodic
scanning).
Thus, traffic flow is represented as continuous flow on theroad network.
The input to the model is the data with respect to road
network represented as the graph with nodes and links, andVS = －α・ s 0 － s β 1 －1
1 + δ・exp(－γ・d )(1)where, VS : acceleration, S0 : desired speed, S : speed of thepreceding car, d : gap distance with the preceding car, α, β,γ, δ : parameters.
The desired speed varies as regulated speed the road, radiusof the curve, and longitudinal gradient.

533

At also varies as driver’s behavior, reflecting the prove car
examination data. Every car can change a lane in responseto the driving situation.

Route choice model into the original traffic simulator withcar-following traffic flow model and fixed route choice

Input of the Desired Speed

Integrated route choice model

Position of the vehicular stop by signal
or Temporary Stop

OD table and road network are made out based on traffic
surveyVehicular Gap and Speed through Car-Following

Execute the simulation to test the reproductivity of traffic
situation

Calculation of the Movement

Execute the simulation on each measures to improve traffic
situation

Fig.2 Calculation of the Vehicular Behavior through Traffic
Flow Model

Fig.3 Examination flow of the traffic simulation
4. THE SUBJECT TO APPLY TRAFFIC SIMULATOR
We can utilize traffic simulator “VISION” in the following
subjects as a tool to estimate the effect and influence of the
project.
4.1 Application to the Measures to Improve Traffic
Congestion
The simulator can be applied to the subject to improve
traffic congestion, such as improving intersections,changing traffic management, and taking grade separation.

Route choice model depends ever-changing travel times oneach road, these are the result of the simulation. The
probability each route is selected takes large it travel time of
the route gets short, applying Dial’s method. Route choice
is refreshed for every time intervals.
Route choice has on significant influence to thereappearance of the simulation result. The higher rankplanning of the project examines the effect and influence tothe surrounding road traffic by traffic assignment procedure
in a wide range road network. To make sure of consistency
with the higher rank planning, the route choice probability
also reflects the result of traffic assignment.

Cars going straight on are
too much

3. PROCEDURE OF THE INVESTIGATIONIn the traffic simulation “VISION”, we integrated route
choice model into the original traffic simulator withcar-following traffic flow model and fixed route choicemodel. Traffic demand, which is expressed as origin anddestination demand matrix, is made out based on traffic
assignment in the higher ranking planning. Also the road
network is made out of a graph with the mass of nodes andlinks. At first, we execute a simulation to test thereappearance of traffic simulation on the present road
network. Traffic volume on each branch in the intersection,
travel speed on each section, length of congestion and so on
is examined comparing the simulated result and observed
values. After the reappearance of the simulation is verified,
we execute the simulation on each measure to improve
traffic situation. The examination procedure is shown in
Figure.3.

Cars turning right affectflows on the main line
Increase of the lane smoothtraffic movement

Providing the lane turningsmooth traffic movement

Fig.4 Example of the Application (Providing Lanes)534

4.2 Application to Road Providing Plannings
The simulator can be applied to road providing planning,such as widening of existing road, providing by-pass andmaking ramps at intersections.

Fig.7 Example of the Application (Improve of the bus stops)
5. EXAMPLE OF THE TRAFFIC SIMULATOR“VISION”
5.1 Estimating the Influence of Railway Elevation
We testified the traffic management measures as thepresumptions to provide station plaza accompanied byrailway elevation project.
Traffic flow not only automobile traffic bat also pedestrians
and bike traffic in the surrounding areas of the plaza wasrepresented precisely.
This simulation made clear the production of new traffic
congestion caused by changing traffic moving as the resultof railway elevating and providing station plaza.

Increase of the lane Improve
traffic movement and air
environment by the roadside

Fig.5 Example of the Application (Increase of the lane)
4.3 Application to Estimate the Influence of the Project
The simulator can be applied to estimate the influence ofthe project, such as changing lanes followed byconstruction work, in and out of working trucks, location ofcommercial facilities and various events.

Fig.8 Traffic Movement by the provision of the Station Plaza
5.2 Estimating the Influence of Traffic Management
Measures on the Arterial Roads
We testified the influence of traffic management measures
to traffic flow in surrounding areas.
The change of travel speed and/or traffic congestion was
represented visually.
In some places of the network, traffic congestion turns hardby introducing traffic management measures. Overall travel
times and emission of carbon dioxide in the objective area
can be decreased by some measures.
Fig.6 Example of the Application (Improve of the cars
going in and out the roadside)

535

1

2

2

4

4

2

2

2

3

下寺町線

1

4

2

1

4

4

4

国道 2 号

2

4

4

駅前幹線

2

4

2

4

2

4
4

4

2

2

2

2

2

内々環状西線

2
船場川線

4

4

凡 例

1
2

1

2

２
1

New
road
新規整備

4

Existing two ways road6

2

現道 対面通行

4

4

2

4

2

Numberof the lane
車線数

4

6

1

2

4

2

2

2

4

2

2
高尾線

4

2

2

4

2

4
4

2
2

3
2

4

3

内環状東線

3
十二所前線

2

3

2 内々環状東線

3

2

1

4

魚町線

Existing one ways road

現道 一方通行

Fig.12 Condition of the Future road Network (about year
2016)
Fig.9 Traffic Movement under Traffic Regulations
6.2 Reappearance of the Simulation
To testify the reappearance of the simulation, 4) traffic
volumes on each branches of the main intersections werecompared with observed values and simulated values. The
correlation coefficient is 0.97, and then we may think ofgood reappearance of the simulation.
Also, the length of traffic congestion and/or travel speed
were compared with observed values and simulated values.
Especially, blocking times and length of congestion were
testified at railway crossings.

6. ESTMATING INFLUENCE TO URBAN AREA
THROUGH PROVIDING NEW ROAD
6.1 Outline of Providing Roads in the Surrounding Areas of
Himeji Station
Now, the project of railway elevation of Japan railway lines,
Sanyo line, Bantan line, and Kishin line, is proceeding. In
December of 2009, railway elevation was finished. And theconstruction of new station building is now under way. In
about 2016, new station building and new station plaza will
be completed. Also, construction work of the roads toconnect north and south area of the station will be finished.
Traffic conditions before and after providing new roads
were simulated to estimate influence to traffic in changing
urban areas.

Fig.13 Traffic Movement through the Simulation on
Existing Road Network

2

2

2

(Vehicles per hour)
1,750

4

2

2

4
下寺町線

3

4

2

4

1

1

1

2

4

4
駅前幹線

国道2号

4

2

4

2

Fig.10 Road Network Surrounding Himeji Station

2

Correlation coefficient = 0.97

2

2
2

2

2

2

4

2

2

4

1

2

2

4

4

内環状東線

1

4
凡 例

1

4

4

2

1

1

2

6

1

2
2

4

2

4

4

6

2

4

2

1,500

4

4
2
内々環状西線

3

2
高尾線

3

2

船場川線

3

Observed traffic volume

十二所前線

2

3

3

2

3

2

1

4

2

魚町線

3

Numberof the lane
車線数
Existing two ways
現道 対面通行
road
Existing one ways
現道 一方通行
road

1,250
1,000
750
500
250

Railway
踏切 crossing

0

Fig.11 Condition of the Existing Road Network (year 2008)

0

250

500

750

1,000 1,250 1,500 1,750

Simula ted tra ffic volume536(Vehicles per hour)

Fig.14 Correlation of the Traffic

6.3 Simulation of the Future Situation
Fig.15 illustrates the future situation of the traffic at thetime when all of the project concerning railway elevation
was completed. The density of roads get intimate by the
project, and traffic flow on the network is made disperseand smooth. Also, congestions at railway crossing arecanceled.

② Benefit of the shortening of Travel Times
Fig.17 illustrates the benefit of the shortening of travel
times. About 25 hundred million yens are generated peryear in the daytime from 7:00AM to 7:00PM.
(million yen per year)

3,000
2,486

2,000

1,000

0

0

Existing

FutureFig.17 Comparison of the shortening of Travel times③ Cutback of Overall Travel DistanceFig.18 shows overall travel times in present and future
situation.
Cutback of overall travel distance is about 7.2% in the peak
hours compared with present situation.
Fig.15 Traffic Movement through the Simulation on FutureRoad Network(vehicles per kilometer)

20,00016,466

6.4 Estimating Effects of the Project(1) Outline of the Influence
Comparing the result of the simulation in present situationand in the future situation, some traffic congestions are
improved considerably, and traffic flow in urban area ismade smooth. Especially, the graphics output as animation
pictures plays an important role in the presentation.
(2) Estimating the Effects
To show the effect of the project quantitatively, thefollowing items were calculated in the present and future
simulation.
① Shortening of Overall Travel Times
Fig.16 illustrates the shortening of overall travel times. It
shows the shortening of overall travel times is about 16% of
the present value in a peak hour.

15,27315,00010,0005,000

0Existing

FutureFig.18 Comparison of Overall Travel Distance④ Cutback of Co2 Emission
Fig.19 Co2 emission shows in present and future situation in
the region.
Cutback of Co2 emission is about 12.8% in the peak hours
compared with present situation.

(vehicles per hour)

1,000

(t)

909

5

765

800

4
600

3.52
3.07

3
400

2
200

1
0
Existing

Future0
Existing

Fig.16 Comparison of Overall Travel Times

FutureFig.19 Comparison of the Carbon Dioxide Emission

537

Traffic management measures produce smooth flow of
traffic in downtown area. That concludes improvement oftransportation in downtown area will reactivate commercial
activities in the area.
Moreover, the open space in downtown area will produce
many activities and spiritual response of human beings.

⑤ Improvement of Average Travel Speed
Fig. 20 shows the comparison of average travel speed on
the main route in the city.
Average travel speed raises 12.5% in the peak hours
compared with present situation.
(km/h)

20

9. REFERENCES[1] T.Akamatsu, “Prediction inducement and control of
traffic flow and theory of dynamic network
assignment”, Infrastructure Planning Review, Vol.18,
December 1995, pp.23-39.
[2] M.Baba, I.Tanahashi, H.Kitaoka, H.Mori and
E.Teramoto, “Traffic simulator NETSTREAM”,
Proceeding of the Society of Information Processing,
Vol.46, January 2005, pp.227.
[3] R.Horiguchi, “Developpment of arterial traffic
assignment model for estimation of traffic
management measures”, Proceedings of Tokyo
University, September 1996, pp.18-21[4] R.Horiguchi, “A consideration on the indexes ofreappearance of dynamic traffic simulation”,
Infrastructure Planning Review, June 2002, pp.1-2.

18
16

10

0
Existing

FutureFig.20 Comparison of Average Travel Speed on theSenbagawa Line
7. POSSIBILITY OF TRAFFIC SIMULATION IN
THE FUTURE
To estimate the effect of transportation projects,
macroscopic models are used in the ordinary analysis
method.
Traffic demand is estimated such that traffic flow is steady.
That means every vehicle is the same and travel speed does
not change on a road for every times. But in actual, every
vehicle is not same and travel time varies by time and byroad conditions. Macroscopic and/or whence a steadymodel has the limit in estimating effects of the projects.
On the other hand, traffic simulation technique has
individual information of the vehicles so. We can get
information taking into account the characteristics ofindividual vehicles. Thus more precise estimation ispossible to the effect of projects.
If dynamic travel demands, i.e. dynamic origin anddestination demand matrix traffic simulation provides
ever-changing traffic flow in the region. Then, we cananalyze traffic conditions and effect of the project more
precisely through the output of visual animations. In the
future, three dimensional computer graphics and/or virtual
reality technique will provide more real and intelligible
pictures of traffic situation. Thus, we will be able to build
more reliable estimating systems of transportation projects.
8. IMPROVEMENT OF TRANSPORTATION AND
LAND DEVELOPPMENT
Improvement of transportation in the urban area in Himejicity, elevation of railway and traffic management measuresconcerned, aims to open the space in the downtown areafrom cars to human beings.
Elevation of railway connects front and back area of thestation in a body. That promotes various kinds of humanactivities in downtown area of the city.
Motorization caused the expansion of urban area anddevelopment in suburban area. That came to subsidence ofthe position in downtown area.
Now, downtown area of the city is being revitalized throughimprovement of transportation in downtown area.

538

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Impact Estimation of Land Cover Changes of Recharge Area on Spring
Discharge in Marshy Environment
Masaaki KONDO and Takamitsu KAJISA, Graduate School of Bioresources, Mie University



ABSTRACT: Many valuable marshy plants, including
carnivorous plants, which were designated as a national protected
species in 1937, have been found in Kanasyouzu. Kanasyouzu is aspring area in Japan’s Mie prefecture situated next to a cliff. The
ground water level is quite high, which makes it possible tonurture wetland plants. In this study, we examined the discharge
characteristics and, parameter estimation of a spring model todetermine how spring water discharge is affected by change to theland cover condition. The annual average discharge wasdetermined to be approximately three liters per second. The springmodel was composed of four tank model and land coverconditions over the top tank. Model parameters were estimated by
optimizer including simulated annealing, which is often used as aprobabilistic procedure, and the Levenberg-Marquardt method,which is not as commonly used. The discharge was calculated
under gravel and asphalt coating conditions that were both better
and worse than the current coating condition. The results showedthat the discharge under the worse condition decreased byapproximately 20 percent and that the discharge under the better
condition increased by approximately 30 percent.

Keywords: spring water, tank model, parameter estimation,
impact of land cover change
1.

INTRODUCTION

Attempts to restore the Kanasyouzu marsh plantcommunity (KMPC), which is located in Mie prefecture,
Japan, have revealed a recent resurgence of marshy
vegetation. Kanasyouzu is an area that includes a spring
situated next to a cliff. According to the local residents'
association, in the past, a lot of water gushed out from
Kanasyouzu - so much that it was used as irrigation water.
KMPC has a high ground water level, which makes it an
ideal environment for many wetland plants. However, land
use has changed as the area has experienced urbanizationand industrialization, and the water discharge has decreased.
After an absence, spring water has recently been identified,and a study of the spring is considered important in order torestore the wetlands.
Tank model was primarily developed by Sugawara [1]. In
terms of spring water analysis, non-linear two-hole tankmodel was developed by Oyama and Mizuno [2]. Kodama
and Komura [3] used four tank model with two outlet of top
tank for spring in Tokyo, Japan. As topographicly-similar
study to this paper, tank model was applied in order toinvestigate hydrological cycle of hilly land where housing
land development had progressed according to [4].
We obtained local annual data relevant to the KMPC,
including rain, air temperature, discharge, and calculated
evaporation in 2010. We then used a four tank model toanalyze the land cover. Optimization, including simulated
annealing and the Levenberg-Marquardt algorithm, wasused to estimate the many parameters needed for the model.

We then examined the effect that changes to the land covercondition had on the spring water discharge.
2.

METHOD

2.1 Location and landscape of KMPC
Fig. 1 shows (a) the location of the Kanasyouzu marsh plantcommunity (KMPC) and (b) a shaded 3D-contour map of
the surrounding area. Most of the slopes around it areoriented northward or eastward. Kanasyouzu is situatednext to a cliff, and the ground gradually slopes up toward
the southwest from the KMPC site. It is located at the footof a mild range of hill. These details helped us to determine
the orientation of restoration area.
The annual amounts of spring water, precipitation andevaporation were used to determine the size of the recharge
area, which is 150,000m2 and indicated by the dotted line in
Fig. 1 (b). The contour map in Fig. 1 (b) was drawn on a 10
meter resolution digital elevation model with fundamental
geospatial data provided by the Geospatial Information
Authority of Japan (Geographical Survey Institute, GSI).
Table 1 lists the area ratio of land-cover classification in therecharge area. There are many factories and companies
related to the automotive industry around the KMPC.
Office buildings, factories, and residences, account for 30%
of recharge area. Gravel parking lots, bare ground, androads account for 25%. Roads and parking lots paved withlow permeability asphalt account for 13%. Sanded school
ground makes up 10% of bare area. Forest, which issituated next to the KMPC, accounts for nine percent. Thereis a small parcel of paddy field in the recharge area.
Non-agricultural land cover comprises to 80% of area of therecharge area.

541

(a)
Fig.1 (a) Location of KMPC marked as star and(b) color-drawn contour map with road (DEM
data provided by GSI was drawn.)

ET

R
f

α
R'
a1

s1

z1
b1
a2

s2

z2
b2
Spring
a3

s3

z3
b3

s4

a4

(a, runoff
coefficient ofside outlet;z, height ofoutlet;b, permeability
coefficient;s, height ofstorage)(b)
Fig.2 Spring modelFig.1 (continued)

Rainfall data was observed by Suzuka Construction
Department which is located 1.3 km away from the KMPC.
The air temperature was determined once in every hour via
an instrument screen located in the KMPC. Possible
evapotranspiration, ETp, was estimated using theThornthwaite system (2) based on monthly averagetemperature [6].

Table 1 Area ratio of land-cover classification and its
run-off coefficientiLand-cover classification

Area ratio
α

1 Building (company19%, house11%)
2 Gravel (parking22%, bare area2%, road1%
3 Asphalt (road10%, parking3%)
4 Farmland
5 Bare area (sand)
6 Forest
7 Paddy field

30%
25%
13%
13%
10%
9%
0%

Runoffcoefficient
f
1.00
0.50
0.80
0.20
0.00
0.30
0.75

ETp j  0.533D0 j (

10T j
I

)a

(2)

T j 1.514)
j 1 512I  (a  0.000000675I 3  0.0000771I 2  0.01792 I  0.49239

Values of f are cited from Tsutsumi et al. (2001).

2.2 Observation method
We set up 90-degree triangular v-notch weir in a channel todetermine the hourly discharge of spring water fromJanuary 27th to December 31st, 2010. We used the weir
because it is useful for measuring small discharges like thespring water in the KMPC. The weir was set at a location
where the water discharged from a paddy field adjacent to
the KMPC did not disturb the measurement of the spring
during irrigation season.
A water-stage recorder was used for the measurement. An
overflow depth of h was determined after the corrections tomake up for the influence of water temperature, barometricpressure, and water level. We convert the formula for flow
by Numachi et al. [5] as

where D0 is possible hours of sunshine in a 12-hour unit
depending on latitude and ETp is possible
evapotranspiration per day (mm/day).
2.3 Analysis
Our spring model was composed of four tank model inseries with land cover conditions over the top tank (Fig. 2).
Precipitation R’ (3) was added to the top tank after taking
into account the permeability rate (1-f) in lieu of R. The
run-off coefficient, f, was treated as constant with time.
Values of f were cited from [7]. We assumed that rain flow
depended on the land cover over the top tank being run off
through street drains beside roads.

R'  R (1  fi )   i(3)i

Q  C  h5/ 2
C  1.354 

(1)
0.004
0.2 h
 (0.14 )(  0.09) 2h
W B

where W is 0.1 m from channel bottom to notch bottom, andB is the 1.5-meter width of the water channel.

where the alpha is the area ratio of land cover in therecharge area.
Possible evapotranspiration, ETp, was estimated by using
the Thornthwaite system [8]. Actual evapotranspiration, ET,
was determined by multiplying the ETp by 0.7. ET was
subtracted from the top tank in the case of a sunny day. In
cases of over 5 mm of rainfall per day, actual

542

0
20
40

40

60
80
100

Rain, mm/day

30

20

10

0

Jan

Feb

Mar

Apr

May

Jun

Jul

Aug

Sep

Oct

Nov

Dec

Time, dayObserved spring

Rain, Vertical lines

Fig.3 Change of spring water and rainfall with timein 20105.0(4)Monthly average discharge, L/s

J   (Qm  Qc ) 2

50

Spring water, L /sec

evapotranspiration was assumed to be zero. Rainfall data
was the observation value. Spring water was assumed to be
the total discharge combined from all outlets beside thetanks.
Analysis of a digital aerial photograph of the recharge areaaround the KMPC enabled us to estimate the area ratio ofland cover, alpha, shown in Table 1 [9].
Considerable skill is required to estimate the many
parameters in this type of tank model. In this study,
simulated annealing (SA) [10] was applied to the model
parameter estimation as a global optimization. In addition
to SA, the Levenberg-Marquardt (LM) algorithm [11] wasused to estimate the parameters because this algorithm has
not been frequently applied as global optimizationaccording to [12], [13]. Fourteen model constants are theunknown parameters. The initial tank depths were designed
for optimizing the parameters. Equation of evaluation
function is as following (4). Time step is one day, and thetotal number of time steps is 339 days.

where, Qm is the measured discharge and Qc is thecalculated discharge.
Model parameters were estimated to use the data between
January 27th and June 30th. These parameters were thenverified from July 1st to December 31st.

4.5
4.0
3.5
3.0
2.5
2.0
1.5
1.0
0.5
0.0

3.

Jan

RESULTS

3.1 Properties of spring water
Fig. 3 shows the changes of the spring water discharge withtime. Rainfall is also shown. From January to February, no
overflow was observed because the rainfall was extremely
low. Several spell of low rainfall resulted in no flooding.
Starting in March, the flooding appeared in response to theincreased precipitation.
The monthly average spring water is shown in Fig. 4. The
average during the winter season ranged from 0.1 to 1.5 L/s,
and was on the decrease due to little rain. In spring season,
the average is 3 to 4.5 L/s, starting out small, and thengradually increasing. The rainfall in July and August in2010 was lower than usual because of the record-breaking,scorching summer. However, between 1 and 4 L/s of springwater was still observed. Spring water at the channel waslast observed in the KMPC in the summer of 2009. During
the fall season, spring water increased due to the heavy
autumnal rain.
The annual average discharge was determined to beapproximately 2.7 liters per second, and the annual amountof spring water was 85,000m3. The spring was notmeasured between January 1st and January 26th because
the amount of spring water was considered to be too little inthe consideration of low rainfall.
We compared the spring water in KMPC with that of aspring near the Kokubunji cliff line because both aresprings near cliffs. A previous report [14] has shown thatthe average spring in the Shinmei-no-mori Mitsuike special
conservation area in Setagaya Ward is 3.2 L/s whennumerically integrated for a times series of spring

Feb

Mar

Apr

May

Jun

Jul

Aug

Sep

Oct

Nov

Dec

Fig.4 Monthly average spring water in 2010
(average indicated with dotted line is 2.7 L/s)
discharges. Although rainfall conditions such as the amountof spring water and rainfall pattern differ every year, the
KMPC was comparable in annual discharge to the Mitsuike
spring famous in Japan.
3.2 Parameter estimation results
The model parameters were estimated with optimizers
including simulated annealing (SA), which is often used asa probabilistic procedure and the Levenberg-Marquardt(LM) algorithm, which is not used as much.
Fig. 5 shows the calibration results of (a) SA and (b) LM.
The measured discharge, Qm, is indicated by a plus markand the calculated discharge, Qc, is indicated by a thicksolid line. While Qc is only passably in agreement with Qm
during the entire term, both agree to some extent early inMarch. It seems that low rainfall during January andFebruary resulted in not only a long dry spell but also thedischarge difference in March. Fig. 6 shows the validation
results of SA and LM. The precious evaluation equation (5),referred to in [15], showed that both methods had equal
accuracy, E = 0.97. In addition, the LM algorithm wasadvantageous in terms of its computation time.

543



E  1  (Q  Q )  100(%)

Q

2

m

c

m

(5)

(b) LM method CAL.

(a) SA method CAL.
0.05

0.05

0

(a)

0

(b)

20

20
40

40
0.04

60

Spring water, m3/sec

Rain, mm/day

Spring water, m3/sec

100
0.03

0.02

100
0.03

0.02

0.01

0.01

0

0

Jan

Feb

Mar

Apr

May

60
80

80

Rain, mm/day

0.04

Jan

Jun

Feb

Mar

Apr

May

Jun

Time, day

Time, dayObserved spring

Calibrated spring

Observed spring

Rain, vertical lines

Calibrated spring

Rain, vertical lines

Fig.5 Parameter calibration with (a) SA method and (b) LM method(b) LM method VER.

(a) SA method VER.
0.05

0.05

0

(a)

0

(b)

20

20
40

40
0.04

60

Spring water, m3/sec

Rain, mm/day

Spring water, m3/sec

100
0.03

0.02

100
0.03

0.02

0.01

0.01

0

0

Jul

Aug

Sep

Oct

Nov

60
80

80

Rain, mm/day

0.04

Jul

Dec

Aug

Sep

Oct

Nov

Dec

Time, day

Time, dayObserved spring

Verified spring

Observed spring

Rain, vertical lines

Verified spring

Rain, vertical lines

Fig.6 Parameter verification with (a) SA method and (b) LM method0.3), the annual amounts of spring water were calculated byusing of the rainfall pattern of 2010 and the model
parameters estimated in section 3.2.
Table 2 shows the annual amount of spring water under acombined land cover. The discharge under the current land
cover was normalized to 100, and it is compared with otherconditions in Table 2. The combination of highpermeability asphalt increased spring water byapproximately 30% compared to the current land cover.
The combination of low run-off coefficients had an effecton the recharge, which resulted in the increased flow fromthe spring.
When the surfaces of parking lots, roads, and bare areas
were covered with low permeability asphalt, the springwater was reduced by approximately 20% compared to thecurrent condition. This means that the worse condition in
low permeability will further reduce the spring water in thewinter season.

Table 2 Impact of change of land-cover condition
on spring water
Low permeable （CURRENT） High permeableasphalt pavementasphalt pavement
Gravel(f = 0.8)(f = 0.3)(f = 0.5)(CURRENT)
Low permeableasphalt
pavement(f = 0.8)
High permeableasphalt pavement(f = 0.3)

4.

80

100
3

111
3

3

(68,373m )(85,722m )(95,312m )

96

115

128

3

(82,287m )3(98,584m )3(109,522m )DISCUSSION

We roughly classified the land cover of the recharge area
into two groups (Table 1). The municipality and companies
take up the roads, parking lots, and bare areas, while
farmland fields and forests are privately owned. Almost allof the recharge area used for non-agricultural purposes is
owned by the municipality or companies. This means thatthe results of this study could have an effect on futureenvironmental strategies.
We examined how changes to the land cover affected thespring. Provided that the asphalt used in roads and parking
lots changed from low (f = 0.8) to high (f = 0.3)
permeability, and that parking lots, roads, and bare areas
made of gravel were turned into such asphalt (f = 0.8, or f =5.

CONCLUSION

We set up a triangular weir to determine the discharge ofspring water in Kanasyouzu. Rainfall in July and August of2010 was lower than usual because of the scorchingsummer. However, even in these months some spring water
was observed. The annual average discharge wasdetermined to be approximately three liters per second,which is equal to that of the famous Mitsuike spring inJapan.
We created a spring model composed of four tank model inseries with land cover conditions over the top tank. The

544

model parameters were estimated by optimizers including
simulated annealing (SA), which is often used as aprobabilistic procedure, and the Levenberg-Marquardt(LM) algorithm, which is not often used. The parameters
calibrated with the LM algorithm were equally as precious
as those with SA. In addition, the LM method wasadvantageous in terms of computation time.
We used the spring model with land cover conditions, andthe estimated model parameters, to calculate the discharge
under both better and worse coating conditions than thecurrent gravel and asphalt have.
The results showed that the discharge under the worse
condition decreased by approximately 20 percent and thatthe discharge under the better condition increased byapproximately 30 percent.
6.

ACKNOWLEDGMENT

The authors would like to extend sincere thanks to Suzuka
Municipal Office for providing the digital aerial
photograph.

REFERENCES[1]

Sugawara M, “Runoff analysis”, Tokyo, KYORITSU SHUPPAN,
1975.
[2] Oyama M and Mizuno A, “Numerical simulation of spring water
discharge (characteristics and application of a nonlinear two-holetank model)”, J. of Groundwater Hydrology, vol.36, 1, 1994,
pp.1-11.
[3] Kodama H and Takamura H, “Water balance of cliff line springwater; For the Seijyo-santyoume Ryokuchi spring water of theTokyo Setagaya Ward”, Chishitsu News, No.655, 2009, pp.50-58.
[4] Ando Y, Mushiake K, and Takahashi Y, "Hydrological cycle of hilly
land and effects of urbanization on it", Proceedings of hydraulicengineering, JSCE, vol.25, 1981, pp.197-208.
[5] Japan Society of Civil Engineers, “Manual of HydraulicEngineering”, Tokyo, Maruzen, 1990, p.284.
[6] Japan Society of Civil Engineers, “Solving Problems based on
Manual of Hydraulic Engineering”, Tokyo, Gihodo shuppan, 1999,
p.37.
[7] Tsutsumi A, Jinno K, and Oheda Y, “Study on the rainwater recharge
model using the groundwater variation”, Proceedings of hydraulicengineering, JSCE, vol.45, 2001, pp.367-372.
[8] Thornthwaite C W, "An Approach toward a Rational Classificationof Climate", Geographical Review, Vol. 38, No. 1. 1948, pp.55-94.
[9] Kondo M and Kajisa T, “Effect of land-cover classification inrecharge area on Spring Changes at Kanasyouzu Marsh Plant
Community”, Proc. of Kyoto branch of JSIDRE, vol.67, 2010,
pp.611-612.
[10] Goffe, Ferrier, and Rogers, “Global Optimization of Statistical
Functions with Simulated Annealing”, J. of Econometrics, vol.60,
1994, pp.65-100.
[11] Doherty J, “Model-independent parameter estimation user
manual:5th edition”, WNC, 2004[12] Kobayashi K, Takara K, and Tachikawa Y, “Parameter estimation ofa distributed rainfall-runoff model by a Levenberg-marquardt
optimization algorithm”, Proc. of hydraulic engineering, JSCE,
vol.51, 2007, pp.409-414.
[13] Kim J H, Paik K R, Lee D R, and Kim H S (2001), "Comparison of
optimization algorithms in parameter calibration of tank model",Electronic version of the XXIX IAHR CONGRESSPROCEEDINGS,<http://www.iahr.org/e-library/beijing_proceedin
gs/Theme_A/COMPARISON%20OF%20OPTIMIZATION%20AL
GORITHMS.html>, browsed on Apr. 1, 2011[14] Tokyo's Setagaya Ward Office (2010) “Report of spring water
investigation at Kokubunji cliff line in Setagaya Ward”, <http://www.city.setagaya.tokyo.jp/030/d00016969.html>, browsed
on Mar. 17, 2011.
[15] Kunimatsu T and Muraoka K, “Model analysis on river pollution”,
Tokyo, Gihodo shuppan, 1994, p.160.

545

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Numerical Study on Characteristics of Levee Breach Flood Disasters in Alluvial
Floodplain
Md. Serazul Islam1 and Tetsuro Tsujimoto2
1

PhD Candidate and 2Professor, Department of Civil Engineering, Nagoya University, Japan.
(Furo-cho, Chikusa-ku, Nagoya, 464-8603, Japan)

ABSTRACT: Levee breach flood disaster is one of the common
natural hazards in Bangladesh. During flooding season, large
amount of water flows through Bangladeshi river and it causes
earthen levee failure and huge damage occurs in the inhabitants
inside the levees. Thus it is important to recognize the process ofinundation and we have attempted to do through numericalsimulation. As for simulation scheme, schematic model area isconsidered with main channel, levee and floodplain, and they areroughly composed of the same sediment characteristics because
the flood plain have been formed by flooding sediment, thelevees have been made by piling up the sediment dredged fromthe river bed. Therefore, we treat these three components
simultaneously in the simulation model.
The main channel, levee, floodplain and flow parameters are
selected in conformity with the study field of Sirajganj district
and Jamuna River in Bangladesh. And RIC-Nays a twodimensional numerical model for flood flow and morphology isutilized in this study upon confirmation through the experimental
and other numerical study. Based on the calculated results,
evolution process of levee breach and inundation of water andsediment in the floodplain are investigated. Levee breach isconsidered to initiate in the middle of the levee with crest
opening. We change the opening size in vertical and longitudinalscales, and the inflow discharge. For the floodplain slope is also
taken into account.

Keywords: Levee, floodplain, sedimentation, inflow discharge.
1. INTRODUCTION
Bangladesh is a disaster prone country where flood andriver embankment failure resulting in inundation andsedimentation to the country side floodplain are commonand frequent natural disasters. On the average, the area
inundated every year is about 26,000 km2, nearly 18% of
the country’s total area [1]. The average flood dischargesof the three main (Ganges-Brahmaputra-Meghna) rivers
individually are within the range 14,000 to 100,000 m3/s
[11]. The mean annual rainfall within Bangladesh varies
from 1250 mm to 5700 mm [13]. During the rainy seasons(June to September), nearly a trillion cubic meter of water
laden with about two billion tons of silt passes through themain rivers, it’s come from upstream catchment in India,
Nepal, Bhutan and China. These sediments settle over theriver beds, floodplains and low lying areas during theinundation and with the recession of flood water. As aresult river channels and their distributaries are silted up
with sediments composed of fine sands and silts causing
drainage congestion and overbank flow resulting in riverembankment failure [8], [9].
Levee breached during the rainy seasons due to excess
load of flooding water. Failure of levees not only increases
the siltation in the river beds and floodplains but it causes

huge damage in agricultural production, residents, roads
and other infrastructures. Due to geographical position of
Bangladesh, we do not avoid levee breach disaster andsuffering to the inhabitants in the floodplain. In order tosave huge amount of money, lives, agriculture and ecology,
it is most important to carry out research on investigationof flood disasters over the alluvial floodplain during levee
breach for the vulnerable area of Sirajganj district inBangladesh. This kind of study is rare in the available
literatures except for few numerical, field investigationand experimental studies which can be found in Tsujimoto,
Mizoguchi and Maeda [14], Aureli and Mignosa [3], andFujita and Tamura [5].
Computational simulation can make comprehensive
analysis on levee breach flooding and its impacts on theentire floodplains. In this paper, we have attempted torecognize the process of inundation and sedimentation in
the floodplain after levee breach in a numerical simulation
scheme for different breach opening lengths, depths anddischarge conditions by changing the inflow duration in
the river channel.
2. DESCRIPTION OF STUDY AREA
Sirajganj district is located in the northwestern part of
Bangladesh. Geographically, extension of Sirajganj
District is within the area of longitude from 89°20’ west to
89°50’ east and in latitude it is 24°00’ south to 24°20’
north. Total area of the district is 2497.92 km2 and it isrelatively a plain land area. Most of the area of this district
goes under water during the rainy season. Total cultivableland is 1799.64 km2, fallow land 157.02 km2, and forestry
0.50 km2. The annual rainfall is 1610 mm. The mighty
Jamuna River is following at the right edge of the district[10]. Fig. 1 depicts the map of study area of Sirajganj
district with Jamuna River and flood protection levee.
Embankment breach is the most common phenomenon for
Jamuna River especially on Jamuna right embankment(JRE). There are so many evidence of breaching ofembankment during severe flood year of 1988, 1998, 2004
and 2007. In the year 2007, JRE at Khokshabari wasbreached of 700 m out of 52 km total length andfloodwater entered into the breached embankment at
Polashpur, Meghai, Dhekuria and Shubhogaccha areas in
Kazipur Upazila of Sirajganj district (The Financial
Express, 27 July 2007). At least 3500 homesteads andabout 6.07 km2 of croplands in 35 villages in the five
Upazilla have damaged in two weeks. The five affected547

Upazilla was Kazipur, Sirajganj
Shahzadpur and Chauhali [7].

Sadar,

changes. Fig. 2 depicts the outline of the simulation steps
for computation.

Belkuchi,Breach at
KhokshabariHydrodynamic
equations

Bay of Bengal

Depth-average
flow fields

Sediment transport
fields

Bed levelchanges

Repeat thesteps

Sediment transport
Equations: Suspended
Initial and boundary (2D advection-diffusion eq.) 2D sediment continuity
conditions
and bed load
equation

Updating time

Updating thecoordinates

Levee
breach/shapealteration

Fig. 2 Outline of model computation steps.
As for simulation scheme, the main channel, levee,
floodplain and flow parameters are selected in conformity
with the study field of Sirajganj district and Jamuna Riverin Bangladesh. Schematic model area is spatially limited
in a part of actual field, which is 5500 times less inhorizontal scale. For the first case of simulation,
computation reach is 900 m long (L) and 500 m wide (Y)(river channel=100 m, levee=25 m and floodplain=375 m)with a bed slope of river channel is 7.5cm/km. Fig. 3 (a, b)
depicts one of the model fields for simulation, whichapproximates the levee breach area of the JRE,
Khokshabari and model layout with computational domain(square cells with ∆x=5 m and ∆y=5 m) of river channel,
breach levee and floodplain, respectively.

Levee
Breach Leveehc mFig. 1 Map of study area.
FloodplainhL =5mh

5m

3. MODEL SET-UP TO DESCRIBE EVOLUTIONPROCESS OF LEVEE BREACH

45m

We analyze a process appearing in river, levee andfloodplain in a same simulation scheme during flood. RICNays, a two-dimensional (2D) model for flood flow andmorphology is utilized in the present study upon
confirmation through the experimental and othernumerical study. It is developed by Hydraulic andHydrology laboratory, Hokkaido University [12]; whereshallow-water equations for 2D unsteady flow expressedin general coordinate system are solved on the boundaryfitted structured grids using the finite-difference method.
Bed-load is calculated by Ashida and Michiue (1972)
equation; the effect of cross-gradient and the influence of
secondary flow, are taken into account [2], [4], [6]. In
considering suspended sediment, an exponential profile ofconcentration is assumed to know planar distribution of
depth-averaged concentration and the 2D advectiondiffusion equations are solved. Finally the bed
deformation is determined using the 2D sedimentcontinuity equation. Equations are solved for the unknown
nodal values by an iterative process. First the flow field iscomputed utilizing initial and boundary conditions; thesediment transport field is then computed, to evaluate
sedimentation rates, and followed by bed topography

37 m

10m 10m
5m
18 m

375 m

Total width, Y= 500 my(a)
Solid Obstacle

500

Y (m)

400
300
200

Floodplain
375 m

Breach Levee
25 m

100

Main channel
0

0

100 m

100 200 300 400 500 600 700 800 900 x

L (m)

(b)
Fig. 3 (a) simulated model area; (b) Plan view of the reachwith river channel, breach levee and floodplain.
The country side bank regions of main channel areprovided with mild slope of 1:9. Levee slope is considered
as Sl= 1:2 on both country side and river side and leveeheight is taken as hL=5 m from the floodplain and 7 mfrom the river bed. Idealized flow and sedimentparameters are considered in the computation: inflow
discharge in the river channel before breach, 750 m3/s (Q)548inflow discharges of 750 m3/s each with duration of 3, 10
and 20 minutes, respectively for levee crest height of 2 mfrom floodplain and floodplain slope is not taken into
consideration are shown in Fig. 5. Initially river water
depth observed high and it decreased by increasing inflow
duration in the main channel because of river flow
diverted to the floodplain through breach with time.
Floodplain inundation depth increased in certain duration.
It can be concluded that after certain period of time at any
specific discharges, inundation depth is not increase in thefloodplain with time. The sediment deposition thickness ishighest at just downstream of the levee breach and it
decreased gradually towards the downstream of thefloodplain with increases distance from the breach location.
Fig. 6 depicts the transverse distribution of velocity andsedimentation profile in the main channel, levee andfloodplain at x =450 m on breach with time. Overtopping
water passes from the river to the floodplain through thebreach initially at high velocity and the velocity reduces
with time because of elevation difference in levee crestand floodplain. River water flows from left to right side in

is considered with the peak flow of Jamuna River inflooding season, and the median size of sediment is chosen
as dm= 0.10 mm for the whole domain. Overflow starts
from the hypothetical notch on top of the levee as a triggerof breach where initial breach is 100 m long (Lb) and crest
height is 2 m (hc) from the floodplain. Though the riverdischarge has a hydrograph in general, a uniformdischarge corresponding to the peak is assumed here. Solidboundary is imposed on the left side of the floodplain.
4. MODEL VALIDATION
The present study model is verified with experimental andnumerical data of Aureli and Mignosa (2001), where they
investigated the flow velocity field near breach in the main
channel without considering morphological change in thefloodplain [3]. They were conducted the experiment in alaboratory flume which was 10 m long and 0.30 m wideand consider the breach region was 2.60 m long and 2.00
m wide. Inflow discharge and slope were maintained equal
to 0.035m3s-1 and 0.1%. The manning’s roughness
coefficient for the bottom and walls of the laboratory
flume was 0.009 and breach width was considered as 0.28
m. The simulation time was 29.5 sec for computed andmeasured velocity profile. Longitudinal velocity
measurement made at different sections with distances of
y = 0.28, 0.26 and 0.16 m in the main channel arecompared with the experimental and numerical values and
Ric-Nays model results which are shown in Fig. 4, thepresent model results agree satisfactorily with theexperimental and numerical data by Aureli and Mignosa
as a whole but it was not perfect match due to the presenceof solid wall normal to the breach opening in a laboratory
flume and it induced strong curvature in the water surface
with a vertical pressure distribution.
1.5
1.25

1

1

0.75

y= 0.24 m
Experimental
Numerical
Ric_Nays model

0.75
1

1.5

1.1

1.2
x (m)

1.3

1.4

1

1.1

1.2
x (m)

1.3

Fig. 5 Simulated flow field in the vicinity of the river,
breach levee and floodplain (S= 0) where hc = 2m fromfloodplain; contours indicate the water depth (h) anddeviation of bed elevation (dzb) with time.

1.4

y= 0.16 mv (m/s)

1.25
1

0.75
1.1

1.2
x (m)

1.3

1.4

Y (m)

1

Fig. 4 Longitudinal velocity (v) profiles comparison
between RIC-Nays model results with Aureli and Mignosa
simulation and experimental data.
5. EFFECT OF FEATURES OF FLOW AND
SEDIMENTATION PATTERNS OVER THE
ALLUVIAL FLOODPLAIN

500
450
400
350
300
250
200
150
100
50
0

3 min
10 min20 min

levee

River
-0.5 0 0.5 1 1.5 2u (m/s)5.1 Effect of Inflow Duration

500
450
400
350
300
250
200
150
100
50
0

Floodplain

y= 0.28 m

Floodplainv (m/s)

1.5
1.25

3 min
10 min20 min

levee

River
-0.5

0
0.5
dzb (m)1

Fig. 6 Transverse distribution of velocity (u) andsedimentation (dzb) profile at x=450 m on breach withtime where hc=2m from floodplain.

In Bangladesh, alluvial floodplain was formed by flooding
and levees were constructed by dredging from river bed.
Normally river bed is lower than the floodplain level. To
observe the time dependent evolution process over thefloodplain due to levee breach of 100 m is carried out forthe main channel and some portion of water diverted to thefloodplain through breach levee. So, it is clearly seen that549

high velocity is observed at short duration near the breachinside the river, top of the levee and as well as floodplain
adjacent to the breach. It is concluded that along the flow
direction velocity is high and it decreased with time. Also,
velocity distribution pattern is not same in the floodplain
due to vortex form in the floodplain. Sediment depositionis occurred near the breach and it decrease along the flow
direction in the floodplain. Depth of sedimentation
increased with time and it is clearly observed at x =450 m.
Fig. 7 shows the comparison of flow and sedimentation
pattern over floodplain where floodplains slope is taken
into consideration as 10 cm/km. This slope criterion istaken from actual field condition. Due to sloping
floodplain, mass transfer rate observed high from river tothe floodplain through breach levee. Velocity of flow ischanged rapidly in compare to without slope condition andslightly more sediment deposited in the sloping floodplain
near to the breach.
Considering small crest opening as 1 m from top of leveeand sloping floodplain as S = 10 cm/km, the transverse
velocity and sedimentation profile in the main channel,
levee and sloping floodplain at x = 450 m on breach withtime are shown in Fig. 8. Inside the river, top of the leveeand floodplain adjacent to the breach velocity is observedhigh and it decreased with time in the main channel. Also,near breach deposition is observed high as larger crest
opening.
Using advection time ratio of 100, the time duration in thesimulation corresponds to 33 hours in the field, which iswithin a possible range and the simulation covers longenough time span.
0.7

0.7

t =10 mint =10 min

0.5

-0.2

without FP slopewith FP slope

-0.5

0.3

without FP slopewith FP slope

0.1
-0.1

125 200 275 350 425 500

Fig. 9 Simulated flow field in the vicinity of the river,
breach levee and floodplain where hc = 2m from floodplain;contours indicate the water depth (h) and deviation of bed
elevation (dzb) with inflow discharges.

125 200 275 350 425 500

YFP (m)YFP (m)

levee

River
-0.5

0

0.5

1

1.5

u (m/s)

500
500
450
450
400
400
700 m3/s
350
350
750 m3/s
300
300
800 m3/s
250
250
200
200
150
150
levee
levee
100
100
50
River
50
River
0
0
-0.25 0 0.25 0.5 0.75 -0.5
00.5
1
u (m/s)dzb (m/s)
Fig. 10 Transverse distribution of velocity (u) andsedimentation (dzb) profile at x=450 m on breach withinflow discharges where hc=2m from floodplain.

3 min
5 min
10 min

Y (m)

500
450
400
350
300
250
200
150
100
50
0

3 min
5 min
10 minFloodplain

500
450
400
350
300
250
200
150
100
50
0

Floodplain

Y (m)Fig. 7 Comparison of transverse distribution of velocity
and sedimentation profile in the floodplain at x= 450 m
on breach considering with and without floodplain slope
where hc=2m from floodplain.

levee

River
-0.5

0
0.5
dzb (m)1

Fig. 8 Transverse distribution of velocity (u) andsedimentation (dzb) profile at x=450 m on breach withtime where hc=4m from floodplain.
5.2 Effect of Inflow Discharges

Floodplain

0.1

Floodplaindzb (m)u (m/s)0.4

Inundation and sedimentation patterns over the floodplain
at different discharges are shown in Fig. 9. To observe thedischarge dependent inundation and sedimentation process
over the floodplain due to levee breach of 100 m is carriedout for 15 minutes each with inflow discharges of 700, 750
and 800 m3/s, respectively. Floodplain inundation depth
increased with increase of inflow discharges in the main
channel. More area of floodplain sedimentation observed
at high inflow discharge in the main channel. Transverse
velocity and sedimentation profile in the main channel,
levee and floodplain at x = 450 m on breach with
discharge are shown in Fig. 10. Adjacent to the breachinside the river, top of levee and floodplain; velocity
observed high at any inflow discharges and then it
decreased suddenly in the floodplain due to river water
passes through the levee crest at a high gradient to a low
gradient in the floodplain. Sediment deposition area isincreased and depth decreased by increasing inflow
discharges in the main channel.

5.3 Effect of Breach Lengths on the Levee

550

carried out for 1 and 20 minutes, respectively anddischarges, 750m3/s and 1550 m3/s for levee crest heightof 2 m and 4 m from floodplain, respectively are shown in
Fig. 13. We have seen that vertical deformation observed
0.4 m in 1 minute at hc=4 m and no vertical erosion at hc=2
m whereas 0.65 m and 0.45 m vertical erosion observed in
20 minutes at hc=4 m and hc=2 m, respectively.

z (m)z_hc = 2m_1 min
z_hc= 4m_0 min
zh_hc = 2m_1 min

z_hc = 2m_0 min
z_hc= 4m_1 min
zh_hc= 4m_1 min

5
4.7
4.4
4.1
3.8
3.5
3.2
2.9
2.6
2.3
2
1.7
1.4

1m
0.4 m

375

400

425

450
L (m)

475

5
4.7
4.4
4.1
3.8
3.5
3.2
2.9
2.6
2.3
2
1.7
1.4

3m

500

525

z_hc= 4m_0 min
z_hc= 4m_20 min
zh_hc= 4m_20 min

1m
0.65 m3m

Crest openingz (m)z_hc = 2m_0 min
z_hc = 2m_20 min
zh_hc = 2m_20 min

Crest opening

Levee breach is considered to initiate in the middle of thelevee with crest opening. To investigate inundation andsedimentation process over the floodplain with different
lengths of levee breaches as 100 m, 200 m and 300 m,
respectively are carried out with duration of 15 minutes
and discharge of 750m3/s which are shown in Fig. 11.
Floodplain inundation depth increased with large breach
opening than smaller one. Sedimentation depth anddeposition area is more when overflow water passes atlarger breach opening. At different breach opening length,
the transverse velocity and sedimentation profile in themain channel, levee and floodplain at x =450 m on breachis shown in Fig. 12. Near breach velocity is observed highat small breach opening and the velocity is reduced byincreasing breach length in the levee. Sedimentdepositions took place over the floodplain along the flow
direction.

0.45 m

375

400

425

450
L (m)

475

500

525

Fig. 13 Time dependent comparison of levee evolution
process with crest opening at top of the levee (y= 115 m).

River-0.5 -0.25 0 0.25 0.5
u (m/s)

z (m)

100 m
200 m
300 m

levee

River
-0.5

0
0.5
dzb (m)(i)h c=2mLb = 100 m
t= 15 minz_original3
z_700 m /s
z_750 m3/s
z_800 m3/s
zh_800 m3/s

350 375 400 425 450 475 500 525 550
L (m)

z (m)

levee

500
450
400
350
300
250
200
150
100
50
0

Floodplain

500
450
400
350
300
250
200
150
100
50
0

Floodplain

Y (m)Fig. 11Simulated flow field in the vicinity of the river,
breach levee and floodplain where hc = 2m fromfloodplain; contours indicate the water depth (h) anddeviation of bed elevation (dzb) with breaching lengths.

5.5
5
4.5
4
3.5
3
2.5
2
1.5

1

Fig. 12 Transverse distribution of velocity (u) andsedimentation (dzb) profile at x=450 m on breach bychanging breach length where hc=2m from floodplain.

6.3
5.7
5.1
4.5
3.9
3.3
2.7
2.1
1.5

(ii)

z_originalz_200 mzh_300 mz_100 mz_300 mh c=2m
Q= 750 m 3 /s
t= 15 min275 325 375 425 475 525 575 625
L (m)Fig. 14 Levee deformation at top of the levee (y= 115 m)
with (i) inflow discharges; and (ii) different breach lengths.

6. LEVEE DEFORMATION DUE TO OVERFLOW
THROUGH BREACH

Not only vertical erosion but also more horizontal erosion
is observed when overtopping water passes through small
crest opening in breach levee at hc=4 m. It can beconcluded that levee breach overflow causes more

Levee deformation is observed along the river flow
direction at top of the levee (y=115 m). To investigate
levee deformation through overflow breach of 100 m is551

hc levee crest heighty distance in y-direction
x distance in x-direction
h depth of flow
zh water surface elevation
dzb deviation of bed elevation
v longitudinal velocity
u transverse velocity
z bed elevation

deformation and consequences as more sedimentation in
the floodplain at small crest opening than larger one.
Obviously, breaching section of higher duration is higher,
and it results that the large amount of discharge causes
serious inundation as well as sedimentation in thefloodplain. To observe levee deformation is carried out for
15 minutes through breaches of 100 m and different
discharges of 700, 750 and 800 m3/s, respectively [Fig. 14(i)] also different initial breach opening as 100, 200 and
300 m and discharge of 750m3/s [Fig. 14 (ii)]. In discharge
dependent conditions (i), slightly more vertical erosion isobserved with inflow discharge increases in the main
channel and breach length dependent conditions (ii),breach enlarged in vertical and horizontal direction for all
length of breaches but vertical enlargement is observed
more when water passing through small breach (100 m)
opening as compared to larger (300 m) one.

8. REFERENCES[1] Ahmed SMU, Hogue MM and Hossain S, “Floods in
Bangladesh: A Hydrological Analysis,” Final Report
R01/92, Institute of Flood Control and Drainage Research(IFCDR), Bangladesh University of Engineering andTechnology(BUET), Dhaka, 1992, pp.1-5.
[2] Ashida K and Michiue M, “Studies on bed loadtransportation for non-uniform sediment and river bed
variation,” Disaster Prevention Research Institute Annuals,
Kyoto University, No 14B, 1972, pp. 259-273 (in Japanese).
[3] Aureli F and Mignosa P, “Comparison between experimental
and numerical results of 2D flows due to levee-breaking,”
XXIX IAHR Congress Proceedings, Theme C, September
16-21, 2001, Beijing, China[4] Engelund F, “Flow and Bed Topography in Channel Bend” J.
of Hydraulic Division, ASCE, Vol.100, No. 11, 1974, pp.
1631‐1648.
[5] Fujita Y and Tamura T, “Enlargement of breaches in flood
levee on alluvial plains,” J. of Natural Disaster Science,Vol. 9, No. 1, 1987, pp. 37-60.
[6] Hasegawa K and Yamaoka S, “The Effect of plane and bed
forms of channels upon the meander development,” J. of
Hydraulic, Coastal and Environmental Engineering, JSCE,
Vol. 29, 1980, pp. 143‐152 (in Japanese).
[7] Hossain MZ and Sakai T, “Severity of Flood Embankments
in Bangladesh and Its Remedial Approach,” AgriculturalEngineering International: the CIGR E-journal, Manuscript
LW 08 004, Vol. X, May 2008.
[8] Islam MZ, Okubo K and Muramoto Y, “Embankment
Failure and Sedimentation over the Flood Plain in
Bangladesh: Field Investigation and Basic Model
Experiments,” J. of Natural Disaster Science, Vol.16, No.1,
1994, pp. 27-53.
[9] Islam MZ, Okubo K, Muramoto Y, and Morikawa H,“Experimental Study on Sedimentation over the Floodplain
due to River Embankment Failure,” Bulletin of the Disaster
Prevention Research Institute, Kyoto University, 44(2),1994, pp. 69-92.
[10] Report on Flood Hazard Model, “Index Based Flood
Insurance Products for Sirajganj District, Bangladesh,:Institute of Water Modeling (IWM), Dhaka, Bangladesh,2010, pp 1-48.
[11] Sarker MH, Huque I, Alam M and Koudstaal R, “Rivers,
chars, char dwellers of Bangladesh,” Int. J. of River BasinManagement, Vol. 1, 2003, pp. 61-80.
[12] Shimizu Y, “Ric-Nays model,” Hydraulic and Hydrology
laboratory, Hokkaido University, Japan, 2009.
[13] Tingsanchali T and Karim MF, “Flood hazard and risk
analysis in the southwest region of Bangladesh,” J. of
Hydrological Process, Vol.19, 2005, pp. 2055-2069.
[14] Tsujimoto T, Mizoguchi A and Maeda A, “Levee breachprocess of a river by overflow erosion,” Int. Conf. of
Fluvial Hydraulics, Portugal 6-8 Sep. 2006.

7. CONCLUSIONSThe conclusions drawn are as follows:i. Floodplain inundation depth is increased in certain
duration. Inundation depth increased with increase of
inflow discharge in the main channel and with large
breach opening than smaller one.
ii. Near breach velocity is observed high both in the riverand in the floodplain at short duration and any inflow
discharges in the main channel. Velocity is reduced byincreasing breach length in the levee.
iii. Sediment deposition is occurred near the breach and it
decrease along the flow direction in the floodplain.
Deposition depth increased with time, discharges in themain channel and breach length in the levee.
iv. Levee breach overflow causes more deformation andconsequences as more sedimentation in the floodplainat small crest opening than larger one.
However, details laboratory experiments and fieldobservation are required to get more clarification about theprocess of levee breach inundation and sedimentation as
well as inhabitant’s suffering who are lived in the alluvial
floodplain. In the study area there are different land use
pattern and effect of flood disasters are not same ineverywhere. In order to get basic differences of flood
disasters using various land type of study area we haveplan for future study.
LIST OF SYMBOLSThe following symbols are used in this paper:
L length of channel
Y wide of channel
YFP wide of floodplain,
SL levee side slope
S floodplain slope
hL height of leveetimet
Q inflow discharge rate
dm mean sediment diameter
length of levee breach
b

m
m
m
m
m
m
m/s
m/s
m

m
m
m
m
min
m3/s
mm
m

552

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Purification System of Ocean Sludge by Activating Microorganisms
Kyoichi OKAMOTO and Kenji HOTTA
Department of Oceanic Architecture and Engineering, CST, Nihon University
Takeshi TOYAMA
Department of Materials and Applied Chemistry, CST, Nihon University
Hideki KOHNO
Department of Applied Molecular Chemistry, College of Industrial Technology, Nihon University

ABSTRACT: Sludge exerts a very big environmental load to local
sea area. Here, attention was paid to micro-bubble technology for
application to the purification of sludge. The important point in thistechnique is to activate the bacteria existing in the area by
micro-bubbles.
In this study, our objects are to develop a new powerful purification
system for sedimentary sludge using a micro-bubble device and by
activating microorganism. As the results of our experiments, we
succeeded in reducing the time needed to purify the sludge.

velocity created by the bubble generating pump.
We used sludge sampled from the ocean, as shown in Fig 3.
Microorganism activator is used; this is liquid and mainlycomprises Kelp and also includes nutrients and someenzymes.
Our used activator is reported to show effective results inpurification for grease trap.

Circulation

Keywords: Micro-bubble, Microorganism, Purification, Sludge

循環給水

1. INTRODUCTION
It is very important to reduce sedimentary sludge in the ocean.
Plans to reduce the sludge are usually dreading or sandcovering. Dredging is a simple way and aims to cut off thesludge. But after cutting off, treating the dredged sludge takes
much more time and, of course, cost. Sand covering, in
general, gives a big load to living organisms and theecological system. Here, a more efficient way is needed toreduce the sludge while not imparting environmental load in
the local sea area.
Now, we have micro-bubble technology. Micro-bubbles canchange conditions into an aerobic state. If the bubbling stops,
the situation changes into anaerobic state, according to recent
research. So, we selected a method for decomposing thesludge by microorganisms.
Therefore, our research targets purification experiments on
oceanic sedimentary sludge by micro-bubbles andmicroorganism activator.

循環
ＭＢ

ポンプ

Micro-Bubble水槽①
Generating

Fig. 1 Experimental Devices

Fig. 2 Experimental Devices

2. EXPERIMENT
2.1 Experimental Devices
The experimental devices consist of two parts, shown in Fig.
1 and 2. The water circulates through two tanks. One tankgenerates micro-bubbles. The micro-bubbles have micro-size
diameter and high solubility.
The other part is the experimental tank. The sludge is put inthis tank. The two tanks are separated due to the high flow
553

Fig.3 Catching Sludge

海水

ヘドロ

水槽②
Experimental
Tank

2.2 Experimental Procedure
At first, the sludge is put in the tank and the micro-bubble
device is powered on. After 6 hours, the microorganism
activator is put in the tank. The measurement items are
dissolved oxygen (DO), water temperature, pH, hydrogensulfide (H2S), total nitrogen (T-N) and total phosphorus (T-P).
In Case 8, only the micro bubble device was operated.
Experimental condition is shown in Table 1.

9.0
8.5
8.0
①
②
③
④

7.5
Table 1 Experimental Conditions
Room
Temp. (degree)

Density of
Activator (ppm)

Circulation
Velocity (cm/s)

①

20

100

0.6

②

23

100

0.6

③

20

200

0.6

④

30

200

0.6

⑤

20

200

5

⑥

20

200

8.4

⑦

20

200

10.7

⑧

20

0

0.6

7.0
0 12 24 36 48 60 72 84 96 108120
time(h)Fig. 5 Changes in pH(mg/l)Case

3. EXPERIMENT
3.1 DO, Water Temperature and pH as Environmental
Conditions for this Experiment
Measured results of water temperature, pH and DO in all
experimental cases are shown in Fig.4 to Fig.6, respectively.
Water temperature up to 12 hours increased rapidly, and thenbecame constant between 25 and 30 centigrade. It seems thiswas caused by the heat from friction of the micro-bubble
device. pH up to 6 hours increased a little and then became
constant. The constant value was 8.0 to 8.6. Dissolved oxygen
up to 6 hours increased a little and then became constant. It
seems oxygen dissolved and reached the saturation point.
From these results, we can recognize anaerobic statechanging into aerobic state.
It seems that these results are very good conditions for growthof microorganisms.

35
30

(℃)

25
20
15
①
②
③
④

10
5
0

⑤
⑥
⑦
⑧

12 24 36 48 60 72 84 96 108120
time(h)Fig. 4 Changes in water Temperature

⑤
⑥
⑦
⑧

10
9
8
7
6
5
4
3
2
1
0

①
②
③
④

⑤
⑥
⑦
⑧

12 24 36 48 60 72 84 96 108120
time(h)Fig. 6 Changes in DO

3.2 Effects on Activator of Microorganisms
Checking the mechanism in this experimental system, thedensity of sulfate was analyzed. Comparing with the effects
on activator of microorganism, experimental conditions areshown in Table 2.
The density of sulfate is shown in Fig. 7. The result was
obtained by the iron chromatography. The slope of thedensity of sulfate in Case 7 is bigger than the one in Case 3.
But the slope in Case 8 is almost no change or less. It seems
that sulfate increases due to the activity of sulfur bacteria. It
also seems sulfate increases according to the circulation
velocity. On the other hand, there is no effect in Case 8 for
only bubbling.
The content of sulfur is shown in Fig. 8. This result was
obtained by the elementary analysis. These are normalized
expression divided by the initial value at time=0. The initialvalues are 3.18, 1.66 and 1.66 in the order of Case 3, 7, 8.
Especially in Case 3, the content of sulfur up to 6 hours did
not change but after 6 hours it decreased rapidly since theactivator of microorganisms was put in the experimental tankafter 6 hours. It seems that microorganisms changed thesludge into the sulfur.
This is similar in Case 8 which is only bubbling. But wecannot see the same thing in Case 7 in which circulation
velocity is fast, and also the values decrease.

554

When the density of activator was 200 (ppm), comparing with
Case 3 and Case 4, the reduction speeds of H2S in Case 3 were
faster than in Case 4, and then the average water temperature
in Case 3 was 25.7 degrees centigrade.
Therefore, the reduction speed of H2S is more effective atwater temperature of about 25 degrees centigrade, but
purification at over 25 degrees centigrade takes a little longer.
Moreover, H2S except in Case 8 was reduced to the lower
limit for measurement within 24 hours. H2S in only bubblingcase in Case 8 took a little longer.

Table 2 Experimental condition for checking the mechanism
Case

Room
Temp. (degree)

Density of
Activator (ppm)

Circulation
Velocity (cm/s)

③

20

200

0.6

⑦

20

200

10.7

⑧

20

0

0.6

2000

(ppm)

③
⑦
⑧

Table.3 Experimental condition for purification effect inrelation to water temperature

1000

0
0

24

48

72

96

Case

Room
Temp. (degree)

Density of
Activator (ppm)

Circulation
Velocity (cm/s)

①

20

100

0.6

②

23

100

0.6

③

20

200

0.6

④

30

200

0.6

⑧

20

0

0.6

120

time(h)Fig. 7 Result of density of sulfate

1(mg/l)

③
⑦
⑧

0.5

0.16
0.14
0.12
0.10
0.08
0.06
0.04
0.02

①
②
③
④
⑧

0

0
0

24

48
72
time(h)96

12 24 36 48 60 72 84 96 108120
time(h)Fig. 9 Changes in H2S

120

Fig. 8 Result of content of sulfur

0.015
0.01
(mg/l/h)

3.3 Effects of Purification in Relation to Water
Temperature
We checked the effects of purification in relation to the watertemperature. Comparing with the water temperature,experimental conditions are shown in Table 3. When thedensity of activator is 100 (ppm), Case 1 and Case 2 areshown. When the density of activator is 200 (ppm), Case 3
and Case 4 are shown. Case 8 is only bubbling by the
micro-bubble device. The H2S is shown in Fig.9, and also the
reduction speed of H2S in Fig.10.
When the density of activator was 100 (ppm), comparing with
Case 1 and Case 2, the reduction speeds of H2S in Case 1 was
faster than in Case 2, and then the average water temperature
in Case 1 was 25.1 degrees centigrade.

①
②
③
④
⑧

0.005
0
0 12 24 36 48 60 72 84 96 108120
time(h)Fig. 10 Changes in speed of H2S

555

8
7
6
5
4
3
2
1
0

allow microorganisms to grow.
Table 4 Experimental condition for purification effect inrelation to circulation velocity
RoomTemp. (degree)

Density of
Activator (ppm)

Circulation
Velocity (cm/s)

③

20

200

0.6

⑤

20

200

5

⑥

20

200

8.4

⑦

20

200

10.7

⑧

20

0

0.6

①
②
③
④
⑧

0.16
③
⑤
⑥
⑦
⑧

0.14
0.12

12 24 36 48 60 72 84 96 108120
time(h)Fig. 11 Changes in T-N

2.5
①
②
③
④
⑧

2.0
(mg/l)

Case(mg/l)(mg/l)

Measured results of T-N and T-P are shown in Fig. 11 and 12,
respectively. T-N has tendency of decreasing. Reduction ratioof T-N has no differences among all cases. T-P has also no
difference among all cases. Here, water temperature does not
influence the reduction ratio on T-N and T-P.

1.5

0.10
0.08
0.06
0.04
0.02
0

12 24 36 48 60 72 84 96 108120
time(h)Fig.13 Changes in H2S

1.0

0.015

0.5
0

(mg/l/h)

0.01
12 24 36 48 60 72 84 96 108120
time(h)

③
⑤
⑥
⑦
⑧

0.005

Fig. 12 Changes in T-P

0
0 12 24 36 48 60 72 84 96 108120
time(h)Fig.14 Changes in speed of H2S(mg/l)

3.4 Effects of Purification in Relation to Circulation
Velocity
We checked the effects of purification in relation to thecirculation velocity. Comparing with the circulation velocity,
experimental conditions are shown in Table 4. Measured
results of H2S are shown in Fig.13 and also the reduction
speed of H2S in Fig.14. The reduction speed in Case 7 was
faster than in the other cases. When the circulation speed
became faster, reduction speed became faster. However, we
could not clarify the maximum circulation speed for suitablepurification. Moreover, reduction of H2S in the case of only
bubbling in Case 8 took a little longer.
Measured results of T-N and T-P are shown in Fig. 15 and 16,
respectively. T-N in all cases decreased. Reduction speed of
T-N in Case 7 was almost the same or a little faster. This wascaused by the high initial value of Case 7. T-P in all cases
decreased. The circulation speed does not influence thereduction of T-N and T-P. T-P in Cases 3, 5 and 6 increased
from 0 to 24 hours, but it did not increase in Cases 7 and 8.
It is assumed the circulation velocity in Case 7 was too fast to

8
7
6
5
4
3
2
1

12 24 36 48 60 72 84 96 108120
time(h)Fig.15 Changes in T-N

556

0

③
⑤
⑥
⑦
⑧

From the comparison with water temperature, it seems
microorganisms are more active at around 25 degrees
centigrade and purification takes a long time when thetemperature is over 25 degrees centigrade.
From the comparison with the circulation velocity, thepurification speed of hydrogen sulfide accelerates ascirculation speed gets faster.

2.5
③
⑤
⑥
⑦
⑧

(mg/l)

2.0
1.5
1.0

5. ACKNOWLEDGMENT

0.5
0The authors would like to express sincere thanks to Mr. M.
Machida, the former graduate student of Nihon University, inJapan.

12 24 36 48 60 72 84 96 108120
time(h)6. REFERENCES

Fig.16 Changes in T-P[1]

4. CONCLUSION[2]

We carried out purification experiments on oceanic
sedimentary sludge by micro-bubbles and microorganism
activator.
As a result, our system could purify the sludge up to 120
hours by microorganism activation. Our system could reduce
the hydrogen sulfide in the sludge to the lower limit formeasurement within 24 hours.
As the mechanism for our experimental system, it seems thatmicroorganisms change the sludge into the sulfur.

[3]

[4]

[5]

557

Hibino, T, and Matsumoto, H (2006). “Distribution of Fluid Mud Layer
in Hiroshima Bay and its Seasonal Variation”, Proc. Civil Engineering,Vol. 62, No 4, pp 348-359.
Okamoto, K, and Hotta, K (2008). “An Experiment on Purification of
Water Quality by Coagulants and Micro Bubble” Book of Recent
Advances in Marine Science and Technology
Mizoguchi, M (1996). “Observation of Water Purification Process in a
Moat by Microbiological Treatment Technique”, Proceedings ofBiological Resource in Mie University, No.16, pp.25-37.
Matsui, R, Okamoto, K, and Hotta, K, (2006). "Water Purification
Experiments by Micro Bubble", Book of Recent Advances in Marine
Science and Technology, pp 119 126.
Matsuo, K, Maeda, K, Ohnari, H, Tsunami, Y, and Ohnari H, (2006).
“Water Purification of a Dam Lake Using Micro Bubble Technology”,
Progress in Multiphase Flow Research I, pp.279-286

Author Index

Fujiwara Yuu
Fumio Tatsuoka
Fumitake Nishimura

A
A.S.M. Abdul Awal
Abdol Hossein Fazli
Abdoullah Namdar
Aiassa Gonzalo
Akihiko Yagi
Akiko Inoue-Kohama
Akinori Morimoto
Akio Tada
Akira Kobayashi
Akira Yuzawa
Aly Ahmed
Amani P.
Angelica Naka
Anna Nakanishi
Apisit Klummeng
Arien Heryansyah
Arrua Pedro
Arthit Neramittagapong
Arwut Yimtae
Ataru SatohAtsushi Koike

211
393
159
221
453
411
527
323
281
517
343
357
333
421
365
407
221
349
203
431
485

G
George Croatto
Giancarlo Flores
Graeme Allinson
Guangqi Chen

Hara T.
Haruka Shimada
Haruki Shimazu
Hasegawa Yuki
Hayano K.
Heitor A.
Heydarnia E.
Hideaki Mizuno
Hideki Kohno
Hidenori Tanaka
Higo Y.
Hiroaki Nakamura
Hiroaki Sato
Hiromu Okazawa
Hirondo Inagaki
Hiroshi Inouye
Hiroshi Sugita
Hirotaka Endo
Hirotoshi Mori
Hiroyasu Ishii
Hiroyuki Ii
Hiroyuki Ii
Hiroyuki Ii
Hiroyuki Ii
Hiroyuki Masuyama
Hisataka Tano
Hormdee D.
Hossain M.B.
Hossain M.Z.

87
163
431
39
485

C
Caiqian Yang
Chaminda Gallage
Chuan Tang
Claverie Eliana

287
197
479
221

D
Dehnavi Y.
Dmitriy Volik

337
77

E
Eberhardt Marcelo

221

177
275
497
299
399
173
357
245
553
317
173
245
123
447
287
533
521
421
97
381
431
457
473
469
97
505
93
191
191, 239, 375,
183

I

F
Fahmida Parvin
Faisal Hj Ali
Faisal Malik
Farsadi T.
Firas A. Salman

431
333
431
479

H

B
Bhuyan Md. Habibullah
Binesh S. M.
Bruce Shelley
Bujang B.K. Huat
Bunei Itoga

57
3

Ikko Ihara
Issam E. Harik
Ito Kazuki
Ito S.

239
167
387
357
167

J

559

275
491
183
177

Jahid M M Islam
Jayantha Kodikara
Jin Sung Ki.
Jing-cai Jiang
Jiraroth Sukolrat
Jiro KuwanoJung Seung Yong.

239
197
227
187
365
87
227

Masayoshi Yamanaka
Masuma Yoshihiro
Matt Kitching
Mayumi Matsumoto
Md. Serazul Islam
Md. Shams Arafat
Md. Toriqul Islam
Md. Zahidul Islam Bhuiyan
Md.Rokunuzzaman
Michiko Masuda
Mikihiro Ioka
Mohamed A. Shahin
Moriguchi Shuji
Moshe Livneh
Mostafa A. Ismail
Motohiro Kawase
Mubarak A. Khan
Muhammad Jamil S.
Mukabi J.N.

K
Kai Zhang
Kaikeerati N.
Kamrul Islam
Kang Wang
Kaoru Kobayashi
Kassai Hirofumi
Kato R.
Katsuo Sasahara
Katsutoshi Suzuki
Kayoko Awaya
Kazuhito Murakami
Kazuo Tani
Kazuya Inoue
Keizo Ugai
Kenichi Fujisawa
Kenichi Horikoshi
Kenji Hotta
Kenta Ochiai
Kentaro Iwashita
Khaled I. Hamza
Kikumoto M.
Kim Tae Ho.
Kimoto S.
Kiyono J.
Kogi S.
Koichiro Otagaki
Koji Sai
Kshama Sundar Roy
Kumara G.H.A.J.J.
Kyoichi Okamoto

187
93
311
485
305
299
15
147
275, 281
501
411
123
275, 281
343
329
381
553
421
287
293
71
227
15, 173425
81, 251
329
329
311
399
553

N
Nagao N.
Nakai T.
Naoki Miyazato
Naoki Sakai
Nasser Arafati
Natsuki Kawabe
Ndeda M.
Noppadol Sangiumsak
Norimitsu Koike
Noriyasu Okazaki
Nuntasarn R.
Ogallo J.
Ogiwara K.
Oka F.
Okado J.
Okajima KenjiÖmer Aydan
Ono Y.

233
479

81
399
15, 17381
141505, 511
425

P
Pairoj Yodsa-nga
Pathmanathan Rajeev
Peng Guo
Pongsakorn Punrattanasin

M
Maekawa Takuya
Maki Oyagi
Makoto Kobayashi
Masaaki Ishida
Masaaki Kondo
Masaaki Takahashi
Masafumi Nakagawa
Masaharu Komamura
Masaki Kitazume
Masanobu Taniguchi
Masanori Ishihara
Masateru Maruyama

71
71
521
147
393
473
81, 251
203, 263, 269
491
323
107

O

L
Lerd Kerdchaiyaphum
Lu Zheng

353
299
431
443
547
311
375
167
299
437
501
117
463
257
117
453
239
387
81, 47, 251

533
453
245
375
541
501
421
447
381
431, 469
97
527

Prinya Chindaprasirt

61
197
501
65, 233, 263,
269, 203
371

Q
Qing Yang

187

R
Rajaraman J.
Rama Mohan Pokhrel
Rezvani R.
Rokonuzzaman M.

560

113
87
337
23

Rusnardi Rahmat Putra

425

Tsuyoshi Baba
Tuan Van Nguyen

S
Saito I.
Sakai R.
Sakai T.
Salehzadeh H.
Sato Shushi
Satoru Ishiguro
Satoru Nakafusa
Satoshi Miyahara
Sawada K.
Sawada Kazuhide
Sekchai Anuvechsirikiat
Seyed Mehdi Mousavi
Shahin H. M.
Shahnazari H.
Shigeko Haruyama
Shinsuke Matsumoto
Shintaro Yao
Shinya Tachibana
Shinya Tsukada
Shouichi Kawabata
Shouta Nakagawa
Shushi Sato
Siau Lian Sang
Sittisak Uparivong
Sutasinee Neramittagapong

71
71
23, 57, 183, 191
337
299
353
305
457
177
463
365
393
71
337
443
329
153
87
517, 521
129, 135
129, 135
329
167
417
349

T
Taira Jinno
Takahiko Nakamura
Takahiro Iwatate
Takamitsu Kajisa
Takashi Hanaoka
Takashi Tsuchida
Takehiro Ohta
Takeshi Katsumi
Takeshi Masui
Takeshi Toyama
Tatemasa Hirata
Tatsuaki Nishigata
Tatsuo Shirakawa
Tawatchai Tho-in
Taylan Z.N.
Tetsuji Okada
Tetsuo Morita
Tetsuro Tsujimoto
Tetsuya Sasaki
Thiruvenkatasamy K.
Toda K.
Tomonori Fujikawa
Tomoyoshi Nishimura
Toru Inui
Toshihiro Morii
Tsuji S.
Tsutomu Tanaka

343
447
505, 511
541
129, 135
129, 135
333
333
153
553
469
153
323
371
31
123
517, 521, 527
547
97
113
71
447
103, 305
333
305
177
275, 281

521
65

V
Vanapalli S.
Vanchai Sata

31
371

W
Wang Liangjie
Wannakul W.
Watcharin Gasaluck
Weimin Gao
Wekesa S.
Wiparat Nisapai

463
107
61
287
81, 251
349

Y
Yamada Toshio
Yange Li
Yasushi Takeuchi
Yoshida M.
Yoshida Y.
Yoshihide Nakagawa
Yoshihito Kogure
Yoshimi Hosokawa
Yota Togashi
Yukiko Saito
Yukimasa Takemoto
Yumie Ito
Yusuke Tomita
Yutaka Toi

299
479
447
177
71
527
521
527
123
97
501
437
153
317

Z
Zhiren Wu
Zhishen Wu
Zhongmin Xu
Zulkifli Yusop

561

287
287
485
407

562

GEOMAT2011 MIE
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTi

ii

PROCEEDINGS OF FIRST INTERNATIONAL CONFERENCE - GEOMAT2011
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT MIE JAPAN, 21-23
NOVEMBER, 2011

Geotechnique, Construction
Materials and Environment
Volume 2
Edited by

Md. Zakaria Hossain
Department of Environmental Science and Technology
Mie University

Toshinori Sakai
Department of Environmental Science and Technology
Mie University

THE GEOMATE INTERNATIONAL SOCIETY, MIE 2011iii

Copyright @ 2011 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrate and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: geomate@gi-j.com
www.gi-j.com

ISBN Number: 978-4-9905958-0-7 C3051

Printed in Japan
iv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Table of Contents

Prefacevi
vii

Organization
Characterization and modeling of various aspects of pre-failure deformation of clayey
geomaterials – applications in modeling
J.N. Mukabi and M.Z. Hossain

563

An emergency medical care information system for fetal monitoring
Muhammad Ibn Ibrahimy

573

Effects of gypsum on the swelling and dispersion behaviour of fine grained soils from Sudan
Abdul Karim, M. Zein and Ahmed M. Elsharief

577

Comparison of bearing capacity equations for vertical central loading
Muhammad Awais Shafique and Tanvir Iqbal Qayyum

583

Pull-out resistance of selected native plants in Hong Kong
Leung F.T.Y., Yan W.M., Tham L.G. and Hau B.C.H.

589

Inland navigation in carrying injured and sick disaster victims in case of a massive earthquake
in the Tokyo metropolitan area
Kazuya Egami, Keiichi Noto, Wataru Miyazaki, Kazukiyo Yamamoto and Takeo Kondo

593

Model calculation and geotechnical interpretation of the correlation co-efficient between finite
and incremental strain path of a formation
J.Rajaraman and K. Thiruvenkatasamy

597

Geocell reinforced bed subjected to simulated traffic loads
P. Sharifi, and S.N. Moghaddas Tafreshi

601

Effects of gradation of various sand deposits of Pakistan on strength of hardened concrete
Gauhar Sabih and S. Muhammad Jamil607

Strengthening of landslide retaining pile wall after geodetic monitoring
Frangov G., Zayakova H., and Hamova M.

613

Response of tapered piles under lateral harmonic vibrations
Ahmad Dehghanpoor and Mahmoud Ghazavi

617

Investigation of ground movement in soil-nailed walls using 1g models and PIV method
H. Mahdavi, and M. Hajialilue-Bonab

623
629

Author Indexv

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Preface

The First International Conference on Geotechnique, Construction Materials and Environment
GEOMAT2011 is organized in Center Palace Miyako Hotel, Tsu city, Mie Japan in conjunction with Japanese
Geotechnical Society (JGS), Japanese Society of Irrigation Drainage and Rural Engineering (JSIDRE), Mie
University, JCK Comp. Ltd and Glorious International. It aims to provide with great opportunities to sharecommon interests on geo-engineering, construction materials, environmental issues, water resources, andearthquake and tsunami disasters. The key objective of this conference is to promote interdisciplinary
research from various regions of the globe. On Friday 11 March at 14:46 Japan Standard Time, the north eastof Japan was severely damaged by the tragic earthquake and tsunami. The conference is dedicated to the
tragic Tohoku-Kanto earthquake and tsunami disasters.
The conference has 3 major themes with 17 specific themes including● Advances in Composite Materials● Computational Mechanics● Foundation and Retaining Walls● Slope Stability● Soil Dynamics● Soil-Structure Interaction● Pavement Technology● Tunnels and Anchors● Site Investigation and Rehabilitation● Ecology and Land Development● Water Resources Planning● Environmental Management● Public Health and Rehabilitation● Earthquake and Tsunami Issues● Safety and Reliability● Geo-Hazard Mitigation● Case History and Practical Experience
The proceedings contain 7 keynote papers along with 115 technical papers from 22 countries. The technicalpapers are selected from the vast number of contributions submitted, after review of the abstracts. The final
papers included in the proceedings have been peer reviewed rigorously and revised as necessary by theauthors. We are grateful to the authors of the contributed papers for their resourceful papers and their help inmaintaining the high assessment of the papers and the co-operation in complying with the requirements of theeditor and the reviewers. We wish to express our sincere thanks to the Organizing Committee Members,
National Advisory Committee Members and International Advisory Committee Members for their valuable
supports. We acknowledge the support of Japanese Geotechnical Society (JGS), Japanese Society of
Irrigation Drainage and Rural Engineering (JSIDRE), Mie University, JCK Comp. Ltd and Glorious
International.

Prof. Dr. Toshinori Sakai
Conference Chairman

Dr. Md. Zakaria Hossain
Conference Secretaryvi

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Organization
Scientific Committees:
Conference Chairman: Prof. Dr. Toshinori Sakai, Mie University, Japan
Conference Honorary Chairman: E/Prof. Dr. Sohji Inoue, Mie University, Japan
Conference Organizing Committee:
Prof. Dr. Takamitsu Kajisa, Mie University, Japan
Prof. Dr. Ishiguro Satoru, Mie University, Japan
Prof. Dr. Sigeko Haruyama, Mie University, Japan
A/Prof. Dr. Atsushi Ishii, Mie University, Japan
A/Prof. Dr. Maasaki Kondo, Mie University, Japan
A/Prof. Dr. Kenji Okajima, Mie University, Japan
A/Prof. Dr. Ryoei Ito, Mie University, Japan
National Advisory Committee:
Prof. Dr. Fumio Tatsuoka, Tokyo University of Science, Japan
Prof. Dr. Kazunori Uchida, Kobe University, Japan
Prof. Dr. Fusao Oka, Kyoto University, Japan
Prof. Dr. Jing-Cai Jiang, University of Tokushima, Japan
Prof. Dr. Akira Murakami, Kyoto University, Japan
E/Prof. Dr. Tadatsugu Tanaka, University of Tokyo, Japan
A/Prof. Dr. Kingshuk Roy, Nihon University, Japan
A/Prof. Dr. Md. Shahin Hossain, Nagoya Institute of Technology, Japan
A/Prof. Dr. Sayuri Kimoto, Kyoto University, Japan
Engr. Hirondo Inagali, President, JCK Com. Ltd. Japan
A/Prof. Dr. Inui Toru, Kyoto University, Japan
A/Prof. Dr. Fujisawa Kazunori, Okayama University, Japan
A/Prof. Dr. Inoue Kazuya, Kobe University, Japan
International Advisory Committee:
Prof Sai Vanapalli, University of Ottawa, Canada
Prof. Dr. Mounir Bouassida, National School of Engineering of Tunis
Prof. Dr. Lilia Robles Austriaco, Angles University Foundation, Philippines
Prof. Dr. A.S.M. Abdul Awal, University Technology Malaysia, Malaysia
Prof. Dr. Muhammad Ibn Ibrahimy, International Islamic University, Malaysia
Prof. Dr. Zahurul Islam, Bangladesh University of Engineering and Technology, Bangladesh
Prof. Dr. Bujang B.K. Huat, University Putra Malaysia, Malaysia
Prof. Dr. Nemy Banthia, University of British Columbia, Canada
Prof. Dr. Ian Jefferson, University of Birmingham, UK
Prof. Dr. John Bolander, Univesity of California, USA
Prof. Dr. Shamshul Chowdhury, Roosevelt University, USA
Prof. Dr. Isabel Pinto, University of Coimbra, Portugal
A/Prof. Mark Jaksa, University of Adelaide, Australia
A/Prof. Dr. William Sebudde Kaggwa, University of Adelaide, Australia
Dr. John N. Mukabi, President of Kenya Geotechncal Society, Kenyavii

Conference Secretariat:
Dr. Zakaria Hossain, Associate Professor
Division of Environmental Science and Technology
Graduate School of Bioresources
Mie University, 1577 Kurima Machiya-cho
Tsu-city, Mie 514-8507, Japan
E-mail: zakaria@bio.mie-u.ac.jp
Tel: +81-59-231-9578、Fax: +81-59-231-9591
Editorial Committee:
Md. Zakaria Hossain
Md. Bellal Hossain
Toshinori Sakai
Md. Toriqul Islam
Executive Committee:
Toshinori Sakai
Md. Zakaria Hossain
Md. Bellal Hossain
Kazuki Ito
Md. Toriqul Islam
Muhammad Ibn Ibrahimy
Ms. Emi Ueno
Ms. Mayumi Matsumotoviii

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Characterization and Modeling of Various Aspects of Pre-failure Deformationof Clayey Geomaterials – Applications in ModelingJohn N. MUKABI* and Zakaria HOSSAIN***EngineerIntel – Research & Development Department, Kensetsu Kaihatsu Consultants, Nairobi, Kenya**Division of Environmental Science and Technology, Graduate School of Bioresources, Mie University, Japanversatile modes of applying the Consolidation and StressRatio (CSSR) functions were also briefly introduced in [1].

ABSTRACT: Having undertaken comprehensive analysis incharacterizing various aspects of clayey geomaterials, theexpanded applications of the fundamental theories andgeo-concepts developed from that knowledge are discussed insimulating the impact of geotechnical changes through modeling.
Of particular interest is the simulation of post-seismic and/or
dynamic loading destructuration. Destructuration andrestructuration models are then introduced through engineering
application of a versatile GECPRO (Geotechnical Engineering
Changes
Probing)
Model,
developedfromamechanistic-empirical geo-mathematical perspective, designed
to probe, predict and simulate changes in geo-properties of clayey
ground and geomaterials. Applications of the GECPRO model inbridging the small strain parameters determined from the field aswell as sophisticated laboratory tests and, the failure parameters
measured in the laboratory, which are vital in geotechnicalengineering design, whilst probing geo-changes, are also
demonstrated.

2. MODELINGPROPERTIESWITH DESTRUCTURATIONASSOCIATED

2.1 Introduction of Generalized Models
The generalized models are proposed for thecharacterization of stiff to very stiff (hard) well cemented
and highly structured Pleistocene clays reconsolidated
tracing any designated anisotropic stress path andsubjected to static and/or dynamic (monotonic and/or
cyclic) loading under undrained and/or drained conditions.
Conceptual models for four aspects of destructuring,
namely; swelling, cyclic prestraining, SHANSEP
reconsolidation and remolding (reconstitution) and twomodels for restructuration (ageing and controlledprestraining) [1], are proposed. The generalized models areschematically depicted in Figs. 2.1, 2.2 and 2.4.

Keywords: model, destructuration, reconsolidation,
restructuration, geo-changes, GECPROM

Overall stress‐strain

1. INTRODUCTION

Intact("undisturbed"
Decreasing
peak shear
strength andDeformation
Resistance

Deviator stress, q (Mpa)

The response and behavior of ground or geotechnicalstructures subjected to seismic loading is, in most cases,
complex and challenging to the geotechnical engineer.
Such challenges can only be progressively solved throughintensive Research & Development involving theapplication of sophisticated testing equipment, advanced
analytical tools, as well as highly undisturbed samples in
the case of laboratory testing. Nevertheless, this can be
economically curtailing particularly for developing andleast developed countries. Furthermore, the continuously
changing physical environment in the face of global
astronomical development and increased intensity andfrequency of seismic action only serves to aggravate thesituation.
Consequently, it is imperative that techniques which areversatile yet reasonably affordable and able to provide
reliable solutions to such challenges be developed.
Mukabi et al. [1] introduced some recently developed
geo-scientific theories, concepts and geo-mathematical
functions providing the basis for some new perceptions
and approaches that can be incorporated within global
geotechnical engineering systems aimed at enhancing
precision and confidence levels of data, parameters, aswell as constitutive and numerical models that are
employed in the characterization of ground andgeomaterials for design, construction control as well as
risk prediction and mitigation. The importance of adopting
the Modified Critical State Theory (MCST) and the mostCyclic prestraining
increasing amplitude andno. of cycles
Degree of
SHANSEP
reconsolidation Increasing(increasing OCR) degree ofswellingRemoulding0

Axial strain, εa (%)

0

Young's Modulus, E or G (Mpa)

Appreciably

Very
SmallYI(b) Stiffness decay model

YI reducing in size
• Decreasing Deformation
Resistance• Reduction in YS and YT sizes
YI: Initial Subyield surface(zone of recoverable linear elasticity)
YS: Secondary Subyield surface(zone of initiation of plastic straining)
YT: Tertiary Subyield surface(Significant plastic straining)
YF: Yield surface at Failure(Large plastic straining)
YyPPF: Primary Post Failure plastic
yielding
YySPF: Secondary Post Failure plastic
yielding

YI tending todiminish

0

Y1I

Y3S

YT5

Y7

SPF
Y9yPPF10 11 12Yy13
14 15 16 17 18 19 20 21

F
Axial
strain, εa (%)
Figure 2.1 Projection Models for: (a) Overall stress ~ strain{(εa)FYS(YF)}., and, (b) Stiffness decay characteristics{(εa)ELS(YI)} ~ {(εa)FYS(YF)}.
0

563

2

4

6

8

Some of the basic mathematical relations that are
applicable in the respective models within YI and YS
sub-yield envelopes are introduced hereafter. Modeling of
the behavior within the rest of the sub-yield can be
determined retrospectively, by interpolation or
extrapolation based on the concepts and equations
introduced in this Study.
It is important to note that, whereas the sizes of the
sub-yield and yield surfaces may be determined from yieldstrain and 3D Transformation Mapping method proposedin this Study, derivation of the shape, directions of strain
increment vectors as well as magnitude and rate ofengagement, engulfment, and diminishing rate is rather
complex [1] and has yet to be investigatedcomprehensively.

2.3 Damage Caused by Prolonged Cyclic Prestraining
The damage and/or change in the geotechnical engineering
properties and characteristics caused by light densification
was investigated by performing Cyclic Prestraining (CP).
Fig. 2.3 compares the stress-strain behavior of lightly
prestrained to non-prestrained characteristic curves. The
results indicate that although the CP effects are virtuallyinsignificant in the YI zone, they become exceedingly
pronounced in the subsequent zones up to post-failure.
Furthermore, the overall stress ~ strain behavior presented
in Fig. 2.3 shows that the CP specimen exhibits aconsiderably lower peak strength and less brittle
characteristics in the region of large scale plastic straining.

2.2 Damage Caused by Swelling
In this case, the simulation considers a model whereby
mainly suction stresses and moisture equilibrium(gradients) change as a result of loading components andenvironmental factors and destructures (ref. to Fig. 2.2).
(a)Pre‐Destructuration Kinematic Multiple Yield Surfaces

YySPF

Figure 2.3 Comparison of cyclically prestrained stress ~strain characteristics for (a) Overall stress ~ strain relations
4% .
{(εa)ELS(YI)} ~ {(εa)FYS(YF)} ε

YyPPF

YF
YT

Cyclic Prestraing structuration/destruction
Model II

Progressive
Destructuration(b)

Initial Yield surface, (εa)YI(Elastic Limit Strain)

YI

YS

Transformation to
Yield Envelopes ofCurve no. 

YySPF
YyPPF
Y : Initial Subyield surface

Reduction

I•Significant
Increase inYF,
YyPPF, and YySPF
Yield Envelopes

YT

YF

YS
YI •Drastic reduction
in YI, YS, and YT
Subyield Envelopes(c) Heavily
DestructuredState(zone of recoverable linear
elasticity)
YS: Secondary Subyield surface(zone of initiation of plasticstraining)
YT: Tertiary Subyield surface(Significant plastic straining)
YF: Yield surface at Failure(Large plastic straining)
YyPPF: Primary Post Failuref plastic
yielding
YySPF: Secondary Post Failuref plastic
yielding

00

αCP Thresholdln

Low αCP

Increasing αCP

Cyclic4 Prestraining
factor,αCP10[function
of
6
8
12
amplitude size, no. of loading cycles & loading…

2

Figure 2.4 Cyclic Prestraining structuration/ destruction
model [Initial Yield Surface {(εa)ELS(YI)}].
The modeling functions are defined as:Transformation to
Yield Envelopes ofCurve no. ln

YyG :ofMerging
Y , Y PPF ,
F

(1)

y

and YySPF

ln

Diminishing

YT
YS

GlobalYyG:Plastic
Yield

(2)

Envelope

YI

Fundamentally, the preliminary study on this topic tends toshow that, depending on the magnitude of the amplitude,number of loading cycles and frequency, CP may enhance
the inherent microstructure linear elastic and recoverable
properties as well as the deformation resistance and shear
strength; or otherwise damage, or even completely destroy

Figure 2.2 Varying states of destructuration models; (a)Pre-destructuration kinematic multiple yield surfaces
depicting mode of engulfment, (b) progressive
destructuration and, (c) Heavily destructured state.
564

the soil structure (microstructure, fabric) of well structured
and cemented Pleistocene clays [1].

behavior of otherwise well cemented and highly structured
Pleistocene clays. This model considers cases whereby
clayey ground is subjected to strong vibrational and/or
earthquake loading, resulting in large plastic straining.
Consequently, characterization of the behavior ofprogressive failure and determination of the original in-situ
properties [2] based on retrospective analysis for purposesof performance based design and prediction, is consideredof utmost importance. A Destructuration Index,
which
defines the degree of destruction as a result of seismic
loading, is introduced as:

2.4 Destruction Caused by Heavy Densification(SHANSEP Reconsolidation)
In this case, destruction caused by heavy densification issimulated by adopting the SHANSEP concept. As
discussed in the preceding sections, densification within
the natural boundary limits of a geomaterial enhances its
properties. However, excessive densification beyond theyield stress leads to large scale straining (softening) anddestruction of cementation, bonding and thixotropic
components as well as causing change in the preferred
particle orientation and inherent/induced anisotropic
properties of natural clayey geomaterials. The basic
definition functions of this characterization are presented
in the following equations where the degree of destruction
due to densification is expressed as a function of OCR.

(7)
where, S f is a soil factor depending on the seismic zone
considering local amplification due to stratification of
subsoil and topographic effects,
is the reference peakground acceleration, is the gravity of acceleration andis introduced intois the soil resistance to deformation.
the CSSR functions [3] that quantitatively correlate thenon-disturbed properties to those of the remolded groundenabling the retrospective derivation of the intact(non-disturbed) initial shear/Young’s modulus and initial
yield strain (Elastic Limit Strain) from the proposed
equations expressed as:(1) For OCR≦2.5

.

10
(Mpa)

(3)

Where, E o R is the resulting initial modulus, E o ref=915Mpa is the reference initial modulus, E o Py is the(8)

pseudo-yield initial modulus determined at the stress level(pseudo-yield stress) which is higher than the yield stress a  y and from which the specimen is rebound [1],is the initial shear modulus of the intact
where,
is the initial shear modulus determinedground andfrom CU TC and CD TC laboratory tests performed on
specimens reconstituted from the original clayey
geomaterial.
The corresponding elastic yield strain is determined as:

expressed as;

E o Py  E o     a Py  ao NCy 

0.39

ao

(4)

and,

=0.225,=1.78,=5.22,=6.85 and= 4.1 are constants for stiff to hard Pleistocene clays
determined in this Study.

(9)(2) For OCR>2.5

E oR  E o  y  OCR
P

3.
 0 . 39

Due to the diagenetic, intricate nature and complexity ofnatural well cemented and highly structured clayey
geomaterials, perfect recoverability is most definitely notenvisaged. Nevertheless, it is considered that, with ageing,
restructuration can be achieved to a certain degree [1]. As aconsequence, the restructuration models are defined
predominantly as functions of secondary consolidation
time. The coupling effect of strain amplitude and loading
rate controlled CP has not been exhaustively examined for
its inclusion in this Study.
The post-destructuration Structural Recoverability Models(SRM) are mathematically defined in the following
equations.

(5)

Deterioration leading to the reduction of the initial yieldstrain {(εa)ELS(YI)} (Elastic Limit Strain) is defined fromthe equation below.

ln

QUASI-RESTRUCTURATION MODELS(6)

Where,

is the resulting size of the initial yield strain,is the initial yield strain determined at the=3.1×10-3 is thepseudo-yield stress level andgeomaterial constant.
2.5 Destruction
CausedbyRemolding(Reconstitution) – Destructuration Model
As discussed in [1], reconstitution of natural clays results
in total remolding and destruction of the vital structuralcomponents, fabric, diagenetic properties, inherent
anisotropy and; transformation to enhanced rheological

ln

(MPa)

(10)

The rate of recoverability for the initial elastic modulus iscomputed from;
565

(11)
The rate of recoverability for the elastic yield strain iscomputed as:(12)
The Secondary Consolidation Time (SCT) required toachieve the structural recoverability initial modulus istherefore computed from:

.

(13)

where,
is the initial modulus after quasi structuralis the post-destructuration initialrecoverability,
modulus determined after Short-Term Consolidation(STC) and=19.3 is LTC related material constant. On
the other hand, the elastic yield strain is computed from,

ln

(%)

GECPROM is primarily designed to be adopted at the
study or investigative stage facilitating basic data andinformation that can then be applied for design offoundations, pavements and other geostructures and/or
comprehensive experimental testing and research regimes,
input
for
globalandconstitutive
models,counter/cross-checking and designating the confidence
levels of laboratory and field experimental testing,construction control, risk analysis and technical mitigation
measures in seismic zones besides the integral role ofprediction of ground response and retrospective behavior.
This model is to be further developed to incorporate the
CSSR modules and MCST concept [1], [3] in order toeffectively complement the ESDAM (Environmental,
Strength and Deformation Analysis Model [4] based onsome of the concepts developed in the process based
sedimentary model that merges the Simsafadim-Clastic
and Discrete Element model [5]. Fig. 4.1 represents thegeneralized web mesh Central Processing (CP) graphicalcomponent of the GECPROM in a 2D p ~q stress plane.

(14)

Hence, the SCT required for SR is then determined as:
.
where

is

(15)
the

elastic

yield

strainafter
is thatquasi-structural recoverability, while=7×10-4. Note that in all
determined within STC and=Δ == .
cases,The degree of recoverability can thence be confirmed fromthe following relations.
ln

(Mpa)
ln

(16)

(%)

(17)

where, E is the post-recovery elastic modulus,
the post-recovery elastic yield strain and=1.4×10-3 are geomaterial constants.
4.

PROPOSED
GEO-CHANGESMODEL (GECPROM)

is=5.42,

Figure 4.1 Graphic representation of the base CPGECPROM module in 2D p q , effective stress plane.
Some of the fundamental functions and mathematicalequations of the GECPROM are briefly introduced herein.
(1) Probing and prediction of shear and elastic moduli
The fundamental equation for the shear modulus
developed from CSSR functions is expressed as:

PROBING(18)

4.1 Derivation of GECPROM Mathematical
Functions Based on Experimental TestingAnappreciably
versatile
mechanistic-empiricalgeo-mathematical model (GECPROM) encompassing
most of the concepts developed in this Study as well asother related studies, is proposed. GECPROM is designed
to probe and estimate changes in vital seismic
geo-properties for clayey geomaterials and ground. The
significant advantage of this model is that various
geotechnical changes and geostructural behavior can bemodeled from a single sophisticated experimental test,whilst simultaneously catering for the effects of drainage
conditions,loading
rate,
andconsolidationstress-strain-time history. The complete and perfect model
would, categorically, have to be multidimensional withrelativistic coaxiality, relations and consequences. The

where
is the initial shear modulus at a variable
stress point ,= ⁄ is the arbitrary or designatedis the initialconsolidation stress ratio traced to ,shear modulus determined at in-situ overburden pressure,=0.95 and
=0.35 are geomaterial constants, thevalues of which are applicable for most natural stiff to hardclayey geomaterials, while =1.16 and =0.4 for stressstates in the 1st quadrant and =-1 for stress states in the 4th
quadrant accordingly.
In developing the model functions it is important toΔ, atconsider the relativistic rates of change ofleast within the 1st quadrant of the { ,q} stress plane for= ⁄ andstress ratio and orientation whereby,566

=and the total derivative is expressed in a
generalized form as:(19)∴

ideally considered to be only stress state dependent, theInitial Yield Strain {(εa)ELS(YI)} representing the thresholdof the initial sub-yield surface, , which is dependent onvarious factors such as consolidation stress-strain-time
history, shear strain rate, drainage conditions and OCR [1],is rather complicated in terms of modeling it in singularity.
In the generalized state therefore, theis expressed
as a function of multi-variables in the form;~(20)
The change of variables within the model is made by
applying Leibnitz’s theorem of the chain rule expressed as:

∑

∙

,

1,2, ⋯ ,,(21)

Whereof, for higher derivatives of products, thedifferentiation generalizes to:
!

∑

!

(22)

!

Furthermore, infinitesimal changes in stress at cross stresspoints are delineated and converged by simply applying
Taylor’s theorem for multi-variable functions expansion;
with as a substitution for

,

,

2

!

Δ

⋯

⋯

(23)

Δ

!

∆ ∙

(24)

Where the deviatoric stress increment magnitude, Δ islarger, exhibiting less error factors, the following equation
may be employed.

ln

,

,

,

,

(26)

,

,

,

,

,

(27)

,

∑

.

(28)

The Boundary Limits (BL) for each of the effects on theinitial yield strain are determined as follows; based on thestationary saddle point of inflection, Taylor’s Theorem ofmultivariable functions (23), Taylor’s Remainder
Theorem and the Euler-Lagrange Equation for stationary
points for Integrals, as well as Maclaurin’s series [6], [7].

The more generalized form applied in analyzing a
multivariable function for stress points sprouting in-directions with X as the vector from the origin, is thenexpressed as;∑,

Furthermore, (27) can be transformed into a more
simplified analytical function whereby the resulting
is considered to be the zonal cumulative magnitude
for each singularity effect, thebounding at thesummation of which is then given by;

Δ

,Δ

,

is the current stress state,
represents thewhere,⁄
isconsolidation stress-strain history,
is thethe secondary consolidation time (ageing),
defines the drainage
overconsolidation ratio factor,
is the strain rate effect. However,conditions andcorroborative deduction made from [1] and the preceding
subsections indicates that the variables can essentially be(ref. to Fig. 4.2).
,,,
simplified as:
The generalized analytical function of the resulting
expressed in compound terms is:

∯∯∮

Δ
Δ

,

I. Stress States ( ) BL
The basic generalized equation defining the impact ofstress states is expressed as:
,

(25)

≦ ,

,

(29)

,

where, constants=0.98,, =0.32, =0.4, for stressstates in the 1st quadrant and, =-1, in the 4th quadrant,
while =1.16.
= ⁄ andthen the Boundary
Considering
Conditions (BC) are defined as:

The model function analyses for (25) are analogous tothose presented in (18) – (24).
(2) Geo-mathematical determination of Initial Yield
Strain (IYS) for varying conditions
GECPROM retrospectively reflects and mirrors theconditions applied during the testing at stress point .
Unlike the initial shear and elastic moduli which can be

0

567

0

30

(εa)ELS DEFINITION MODEL

Δ

lim

Δ

Δ

ANALYTICAL FUNCTION MODEL

0

→∞

lim

1

3
Δ
Δ 3

0,

3
Δ
Δ 3

0

→∞

0

Local Saddle Pointof Inflection

Linear‐elastic recoverable
geoproperties

Δ
Δ

Δ
3

Δ

0

Δ 2

2

Δ 1

0

Δ 3

4

Δ 4

Δ5,Δ4

Δ 5

5

1

Δ n 4 ,Δ n 3

2

Δ 11

Δ1

Δ n 3 ,Δ n 2

Δ4

3

Δ2

Δ

4

1
4

Δ 12

5

Δ3

Δ5

Δ

1
5

Δ 31

Virgin Loading
Unloading
Reloading

Δ n 2 ,Δ n 1

Δ n 1 ,0Δ

1Figure 4.2 Resulting Initial Yield Strain,{(εa)ELS(YI)} simplified Analytical Function Model (AFM)

and;

∑ ∑

0

0

(31)

Carrying out 2nd order partial differentiation w.r.t
obtain,Δ

Defining a matrixΔhaving elements given by
,

we

(35)

(36)

Equation (35) can then be rewritten in the form;
1
0

(32)

Performing the same w.r.t| |Δ

we obtain,

Δ| |

(37)

where | | is the column vector with Δ
as itscomponents and | | is its transpose. Since is realand orthogonal
and symmetric it has eigenvalues
eigenvectors , which can be normalized to the form;
,

,

(38)

where the Kronecker delta symbol
1 for
and
0 for
. Expanding | | in these terms yields;

1

| |

∑

,

(39)

where,are coefficients dependent upon
Substituting this into (41) yields;
0

(33)

∑

(40)

∑
minimum; Δ
0 : for all sets ofvalues of the∑
0 :therefore,(ii) maximum; Δeigenvalues of >0∑(iii) saddle; Δ
0: therefore, eigenvaluesof >0Considering (34), which is a two variable function, i.e.
= ⁄ andthen the matrix is given by

(i)

!
→0

Δ| |

Consequently, for the stationary point to be;, indicated in the lower BL of the second
component, which defines the interface of the transition
to theand transposition from the end of
depicted in Fig. 4.2 is solved
beginning of
analogous to (23) along with Taylor’s Remainder Theorem
resulting in the function;, |lim →

| |

Δ

| | .

(34)

In the case where it is necessary to determine the nature ofa stationary point and stationary values for multivariable
functions, the function is expanded as a Taylor series about
the stress point in reference to (24), which is the full formof the approximation;(41)

Hence,

Δ
568

(42)

V. Strain ratelnwhich is solved to yield;

where, the subscripts SR denote Strain Rate, ASR: Applied
Strain Rate during testing or arbitrarily designated and
RSR: Reference Strain Rate.

2

(43)
4
Equation (43) is then solved to determine the minima.
An analogous procedure is applied for analyzing thefunctions of the other controlling factors whose
generalized equations are presented as follows.
II. Consolidation Stress-strain history (

)

ln

(44)

where
=1.9×10-3 is a constant,is thesecondaryconsolidation
factor,istheis the initial yieldoverconsolidation factor andstrain determined under normally consolidated conditions
at a standard time period designated after the end ofprimary consolidation.
III. Drainage conditions (where, Δ = Δ

where Δ , Δ are the effective axial and radial stresses,
respectively determined at the threshold of

Δ

(50)

(46)

Δ

(51)

Substituting (49) into (50) and expanding as a Taylor series
in , then the following equation is obtained.
Δ ,

Δ

,Δ

,Δ

Δ

)

IV. Cyclic prestraining (

ln
,

→Δ Δ

0

where
,
are the initial drained and undrainedelastic moduli respectively.

0is the function required to makewhere, parameter is small and Δ
is an arbitrary
function with sufficiently amenable mathematicalproperties. For the value of to be stationary w.r.t suchvariations,is the stress ratio during consolidation,=-1,=+1 ( : drained and : undrained) and is
the Poisson’s ratio. For perfectly drained conditions ≑0.2
and ≑0.5 for perfectly undrained state. Under partiallycan be determined as
undrained conditions

1

(49)

/ Δ .

Suppose that Δ Δ
stationary then,
Δ Δ

⁄

2

Δ

Δ ,Δ ,Δ

2 Δ
Δ
(45)

1

4.2 Examination of Abrupt Variations in Small
Stress-strain Characteristics
In cases whereby the stress-strain characteristics change
abruptly as a result of, for example, excessive cyclicprestraining, swelling, excessive densification or otherform of destructuration, structuration or geotechnicalchanges, the theories and concepts related to the calculusof variations are applied [6]. In particular, theEuler-Lagrange (EL) equation is applied provided that allfunctions are sufficiently smooth, differentiable and withindifferentiable spaces.
Consider that the characteristics in Fig. 4.2 change
drastically and as a consequence, the functions defined in
the preceding section can no longer satisfy any of theconditions within the Δ ~Δ space. It would then be amatter of extreme importance that we are able to evaluate
and determine the forms of such changes in order to retrofit
the analytical functions.
Take for example that stress~strain characteristic changes
occur within the range
andthen, examining
the changes on the basis of the EL approach yields;

)

Δ
Δ

2

(48)

3
3

,
0

Δ ,Δ ,Δ(47)

Δ
With this form, for Δ ,that for all values of Δ

Refer to Fig. 2.3 depicting cyclic prestraining
structuration/ destructuration model and the modeling
functions.

569

Δ
0

(52)

the condition in (51) implies, we require;

Δ

0

2 1

(53)

where, denotes the first-order variation in the value ofdue to the variation defined in (49) for the function
Δ Δ . Integrating the 2nd term by parts yields;

1
0

(59)
2
1

≦ 0.1 and,

0≦

0.1

and,

0(54)

Δ

0

10

%

(55)

When >0.1(1) For Intact specimens

Consequently, since (53) must be satisfied for an arbitrary
function; Δ , then;10(MPa)(56)(2) For Reconstituted specimens

Other Euler’s equations applied take the form:
(57)



Δ

0

(58)

and ∂Δ∂.

(62)

Consequently, the Δ ~Δ relations are generated throughthe integration of (59) ~ (62), i.e.;

and,

For 0≦ ≦0.1%
Δ

where, ∂Δ

∂

,Δ

(61)

10

Δ

Δ

(60)

and
are determined simultaneously atwhere,=0.001% and 0.01%, andis derived from (66).

,and, Δ,
ConsideringΔ
are fixed, then it becomes a prerequisite
0, in which case;condition that;

,

.

Δ

. %

4.3 Application of GECPROM in Modeling ofPre-failure Deformation of Stiff Pleistocene Clays
Adopting the GECPROM, pre-failure stress path,stress-strain and stiffness characteristics are modeled.
Typical comparisons of the modeled vs. experimental
curves are presented in Figs. 4.3, 4.4 and 4.5, respectively.



Δ
10 Δ

For

(63)

>0.1%
. %

Δ

.

Δ

.

10 Δ

10 Δ

(64)

(65)

The sub-yield surface strain limits are mathematicallydefined as:Initial elastic yield surface strain limit,
⁄

0.1
,

,

%

(66)
0.1

.

⁄

6
,

Figure 4.3 Comparison of experimental and modeled Shear
Stress Paths (SSP)
where
The paraphrased integral model equations applied in the
GECPROM, along with others not reported in this paper,for generating the simulation and prediction curves, areexpressed in a generalized form as:

,

%

(67)

then,
,


570

Secondary sub-yield surface strain limit,

(68)

.

.

.

(69)

.



Tertiary sub-yield surface strain limit,
.

.

.

.

%

.



(70)

Ultimate yield surface strain limit,
.

.

%

'

ε

‐ .

'

‐A ln OCR

'

(74)

'

On the other hand, at any given Stress State (SS), theLimiting Elastic Modulus,
is computed as follows.

10⁄= 1 for HOC andiswhere,
the elastic modulus determined at the in-situ overburdenpressure (), while the limiting initial yield strain,
can be retrospectively extrapolated by adopting
(74).

(71)

(3) Limiting Elastic Modulus (LEM) at given SS

0.1

(73)

(72)

This represents the Limiting Elastic State [LES], beyond
which the elastic and/or shear moduli cannot increase
notwithstanding any further effects such as OCR, ageing,strain rate, cyclic prestraining, and drainage (which would,
on the contrary, be detrimental beyond this state) provided
that the soil fabric does not undergo change and is devoidof extrinsic geomaterial influence and/or contamination.

where,=3.1×10-3 for intact specimen and 1.52×10-3 forreconstituted specimens are constants applicable for stiff tohard clayey geomaterials.
The LYS is vital in determining the ultimate Boundary
Surface (BS) and also establishing control parameters
applicable in design, risk analysis and construction control.
The results observed in Figs. 4.3 ~ 4.6 show virtuallysimilar characteristics between the modeled andexperimental curves for the very stiff, well structured andoverconsolidated OAP Pleistocene clay, thereby verifying
the confidence level of the GECPROM at this stage.

(4) Limiting Yield Stress (LYS)
equation can
The Limiting Yield Stress,therefore be derived from (4) ~ (6) and (72) as:

Figure 4.4 Comparison of modeled and measured stress~strain relations for Over-Consolidated (OC) OAP clay for: (a)1% ; (b) small strains {(εa)SYS(YS)}, ε
0.01% and, (c) very
intermediate strains {(εa)ELS(YI)} ~ {(εa)TYS(YT)}, ε
0.003% .
small strains {(εa)ELS(YI)}, εFigure 4.5 Comparison of modeled and experimentally determined strain level dependency of stiffness decay curves
1% for ageing effects for: (a) intact and reconstituted
including predicted curves {(εa)ELS(YI)} ~ {(εa)FYS(YF)}, ε
1% and , with prediction for ageing for: (b) intact and reconstituted
specimens, {(εa)ELS(YI)} ~ {(εa)FYS(YF)} for ε
specimens and, (c) intact specimens only.
571

relativistic consequences. This is certainly a subject for
further research. However, it can derivatively beconcluded that:
1) The methods of quantitatively determining themagnitudes of the zonal sub-yield strain and theLimiting (Critical) Elastic State proposed in this Study
are useful in modeling and prediction of geo-structural
behavior, risk analysis and construction control.
2) The proposed GECPROM is versatile and appreciably
effective in probing, simulating, modeling andpredicting geotechnical changes in ground andgeomaterial properties.
3) The GECPROM mathematical functions and modules
may be useful in correcting the models that have beendeveloped based on tests performed on reconstituted
clays, in order to model the behavior of natural clays.

On the other hand, Fig. 4.6 is a graphical depiction of thecomparison of modeled and experimentally derived
multiplekinematic
sub-yield
surfaces;
hypothetic-empirically bound around the respective
computed Yield Strain Limit [{YSL {(εa)YLS(YLS)}] related
congruent p’~ q stresses.
The following derivations can be made from this figure: 1)
the intact specimen exhibits much larger sub-yield as wellas yield surfaces compared to the reconstituted one; amanifestation of destructuration effects; 2) the orientationsof reconstituted specimen surfaces are different fromthose of intact surfaces in all cases; a manifestation of thedifferences in structure; 3) by extension, it can therefore beconsidered that the intact and reconstituted surface sizes
vary in all directions; 4) a link exists between themagnitude of stiffness and the configuration of themultiple kinematic sub-yield surfaces; 5) within the Yu
surface, the magnitude and stress path rotation increases as
the size of the sub-yield surfaces becomes larger; 6) the
farther the distance of the current stress point from thefailure line, the closer the range of engagement of the
neigboring surfaces and, 7) the modeled and experimetal
curves are practically the same in all cases demonstrating
the confindence and reliability levels of the GECPROM.

6. ACKNOWLEDGMENT
The authors are highly indebted to Prof. Fumio Tatsuoka,
formerly of the University of Tokyo. Special appreciation
is extended to the Japan International Cooperation Agency(JICA), Taisei Corporation, Japan, Kajima Corporation,
Japan, for sponsoring most of this research and the supportof the Materials Testing and Research Department,
Ministry of Roads and Public Works, Kenya.
REFERENCES[1]

Mukabi JN, “Characterization and modeling of various aspects ofpre-failure deformation of clayey Geomaterials – Fundamental
theories and analyses”, Proc. 1st Int. Conf. on GEOMAT, Mie,2011.
[2] Tatsuoka F, “Laboratory stress-strain tests for developments ingeotechnical engineering research and practice”, Bishop Lecture, tobe published in Proc. Int. Symp., Seoul, 2011.
[3] Mukabi JN, Kotheki S, Mathematical derivative of the modified
Critical State Theory and its application in Soil Mechanics. Procs.
2nd International Conf. on Applied Physics & Mathematics, 2010
IACSIT, Kuala Lumpur, pp. 484-492.
[4] Mukabi JN, Kotheki S, “A Geo-mathematical model for
quantitative analysis and prediction of the impact of environmental
changes on the characteristics of geomaterial formation”, inProceedings of 14th Int. Conf. on Mathematical Geosciences,
Budapest, 2010, pp 382-406.
[5] Carmona A, Clavera-Gispert R, Gratacos O, Hardy S, “Modelingsyntectonic sedimentation: Combining a Discrete Element Model oftechtonic deformation and a process-based sedimentary model in
3D”, J. of Mathematical Geosciences, vol. 42, No. 5, July 2010, pp.
519-534.
[6] Riley KF, Hobson MP, Bence SJ, Mathematical Methods forphysics and engineering .Cambridge: Cambridge University Press,
1998.
[7] Dass HK, Advanced Engineering Mathematics. New Delhi: S.
Chand & Company Ltd., 1999.
[8] Jardine RJ, “Investigations of pile-soil behavior with specialreference to the foundations of offshore structures”, Ph.D thesis,
Imperial College, University of London, 1985.
[9] Jardine RJ. “Some observations on the kinematic nature of soil
stiffness”, Soils & Foundations, JSFE Vol. 32, No. 2, 1992,
pp111-124.
[10] Atkison JH, Stallebrass SE, “Experimental determination ofstress-strain-time characteristics in laboratory and in-situ tests” inProc. of 10th Eur. Conf. on SMGE. Florence 3, 1991, pp. 915-956.
[11] Stallebrass SE, Taylor RN, “The development of a constitutive
model for the prediction of ground movements in overconsolidated
clay”, Geotechnique, vol. 47, No. 2, 1997, pp. 235-253.

Figure 4.6 Comparison of modeled and experimentallyderivedmultiple kinematic sub-yieldsurfaces;
hypothetic-empirically bound around the Yield Strain
Limit [{YSL {(εa)YLS(YLS)}] congruent p’~ q stresses.
As hypothetically suggested by various researchers; [8] –[12], the scientifically determined results of this Study
show that the history surface, YT is highly susceptible to therecent stress history and that the memory of the previous
history is temporarily maintained notwithstanding changes
in loading direction and stress path (also ref. to [1]).
5.

CONCLUSIONS

Although these results confirm some theories, hypothesis
and concepts that have been adopted in most constitutive
multiple kinematic hardening models such as thetranslation laws, magnitude of strain development andmode of changes in stiffness, they also seem to contradict
some others such as the dragging of surfaces by current
stress states, their geometry, size, configuration and[12] Tatsuoka F, Jardine RJ, Presti DL, Benedetto HD, Kodaka T,“Characterizing the pre-failure deformation of geomaterials”. XIV
IC on SMFE, Theme Lecture, Hamburg, 1999, pp. 2129‐2164

572

First International Conference on Geotechnique, Construction Materials
and Environment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7

An Emergency Medical Care Information System for Fetal Monitoring
Muhammad Ibn Ibrahimy
Department of Electrical and Computer Engineering
International Islamic University Malaysia
Kuala Lumpur, Malaysia

ABSTRACT: Developing and less developing countries are
identified in risk of stillbirth due to insufficient emergency medical
take-care supports. The regular check-up of the fetal
electrocardiography could be an earliest prevention of this occurrence.
This paper presents a research work aims to implement a simple,
customize and portable emergency medical care system for fetal ECG
monitoring from remote location. After acquisitions of the ECG
signal from the abdomen wall of a pregnant woman (patient), thelocal terminal then transfers this ECG signal as data to the remoteterminal at the expert physician for diagnosis purpose. At the remoteterminal, the fetal ECG signals are then extracted from the ECG
signals. In this case, computer network establishes a bridge between
the patient and a distant expert physician. The network program
developed based on the client/server applications. The developed
network program is capable to support both the ECG data transfer
and online chat session simultaneously. A specialist physician at the
remote terminal can diagnose the fetal ECG signal and provides
instruction to the local terminal in case of emergency. A number ofcases studied by this developed system and found approximately
same result compared with a commercial one.

Keywords- Telemedicine; ECG; AFECG; FECG; FHR
1. INTRODUCTION
Nowadays, telemedicine is producing a great impact in ourdaily life for health take-care support, especially the patient
whose are located in remote nonclinical environment or if
there is no expert physician. The stillbirth rate is still higher inthe developing and less developing countries compared to thedeveloped countries due to insufficient emergency medical
take-care supports. The telemedicine support for the FECG
monitoring systems can be a sort of solution of this crisis. An
FECG monitoring systems based on the telemedicine supportconsist of three major parts: (1) data acquisition system, (2)networking system, and (3) fetal hart rate (FHR) monitoring
system. The AFECG data acquisition system acquires the fetal
ECG signal from mother’s abdominal wall and stores into a
local computer terminal as a data. The network application
serves the purpose to transfer the AFECG data to the remoteterminal for diagnosis purpose. The AFECG signals are thenprocesses at the remote terminal and extract the FECG fromthe AFECG signal by an FHR monitoring program. However,still the total systems are huge, complex and costly for
individual. The AFECG signal obtains from the abdominal
wall of the pregnant mother contains many vital information
regarding the fetus. This signal provides significant
information about the heart rate, arterial blood oxygenation,
blood pressure and respiratory-rate, etc [1]. Most of thestillbirth occurs during the maternal sleep, since during thisperiod, blood pressure of the mother becomes lowest, and alsoa cord compression occurs when the mother is lying down
especially, in late term while there is a little space for the baby
to move. An FECG monitor systems may have a potential
effect to detect any dangerous fluctuation or decline in the
baby's ECG and allows some scope for appropriate medical
intervention. To observe such abnormalities, compact,
portable, long-term ambulatory monitoring system hasapproved by the mainstay of fetal surveillance during
pregnancy [1].
2. METHODOLOGYThis research has been divided into three parts such as (1)developing a compact portable AFECG data acquisition
system, (2) developing a customized network program forcommunication between local terminal and remote terminal(3) improving the accuracy of the existing techniques forautomatic analysis of the AFECG signals for FHR monitoring.
The overall flow of the research work as shown in Fig. 1.

Figure 1. The basic block diagram of the overall system.

2.1 Data Acquisition System
The purpose of the data acquisition system is to collects thematernal AFECG signal and sends it to a local computer
through RS232 serial port. The system developed based on acommercial ECG front-end ASIC hardware module named“CARDIC”. A microcontroller module (18F4550) controls all
the functions of the system. An optical isolator is used in
between the system and computer for electrical isolation,
which prevent the damage from unexpected electrical shortcircuit condition as shown in Fig. 2.

573

is successfully setup, the user can send the data from local
terminal to the remort terminal. The physician at the remotelocation can display the fetal ECG graph with the graph plot
features of the RPMS. In addition, both the users cancommunicate via a simple chat program. The Fig. 3(a) and Fig.
3(b) are the screenshot of LPMS and RPMS respectively.
2.4 Summary of the Network Program
Local Patient Monitoring System (LPMS) provides:

Dial up to predefined hospital or clinic terminal andhence setup a full duplex communication with RPMS
Setup a chat/dialog session with RPMS
Transfers the AFECG data

Remote Patient Monitoring System (RPMS) provides:

Figure 2. Pictorial view of the AFECG data acquisition systems

2.2 Networking
The networking part is developed by using Visual Basicclient/server pair applications. In brief, this part of the projectis included two sections (1) chat section, and (2) file transfer
section. The main purpose of the chat section is to establish aconnection between patient (client) and physician (server), sothey can make conversation to each other on real time. On theother hand, the file transfer section establishes a network
connection using TCP/IP protocol, so that client can transfer
any types and any size of data file(s) to the server terminal.
The computers are identified by their 32-bit IP address and a
16-bit number is used to identify the ports during the network
communication.

Setup a network connection with LPMS
Setup a chat/dialog session with LPMS
Receives the AFECG data and processed for graphplots

2.5 Networking Module and Algorithm

2.3 Software Design and Development
The network program is in the form of client/server pairapplications. The client application, known as Local Patient
Monitoring System (LPMS), which is installed in the local
terminal.








(a)

(b)


(a)

Figure 4. (a) LPMS networking module flowchart and (b) RPMS
networking module flowchart.

(b)Figure 3. (a) Local Patient Monitoring System (LPMS) and (b) Remote
Patient Monitoring System (RPMS)

The network module of LPMS works in synchronization with
its counterpart RPMS. The function of these modules is tosetup a full duplex connection between the local terminal andthe remote terminal [2]. Multithread programming concepts
are used for developing the network modules. Multithreadprogramming is necessary due to the requirement ofmultitasking operation [3] in the networking applications. For

On the other hand, the server application, known as Remote
Patient Monitoring System (RPMS), is deeds to be install in
the remote terminal. The most fundamental feature of the
network program is to setup a connection between the local
and remote terminal via TCP/IP. When a network connection574

file-sending algorithm except, in this case data will be readfrom the I/O stream. The flowchart of the file receivealgorithm is shown in Figure 5b. The socket and I/O stream
will remain open until the RPMS exited.

instance, some times LPMS is required to hold a chat session,at the same time it sends the AFECG signal RPMS terminal.
Figure 4a and Figure 4b are the flowcharts for the LPMS and
RPMS respectively. At any instance, when the client isconnected to the X number port of the server, then the server
assigned a thread that waits for its connections at port X. Ones
the connection is accepted, the client and the server cancommunicate with each other through the connection bound at
the port X. file module (in LPMS)

Chat Session Module Algorithm:
The chat session module is intended to provides connection
for communication between the patient (or patient’s attendant)
at LPMS and physician at RPMS. This is necessary because
the physician cannot know the patient’s all health condition
only through the data file, but also need to know the healthcondition during some discussions. Moreover in case ofreceived data or file is corrupted, the physician can request for
retransmission. Also a chat session can enable the online
consultation.

Sending/Receiving Data Module Algorithm:
The send file and the receive file modules in the LPMS andthe RPMS respectively are responsible for enables thetransmission and reception of AFECG data via the networking
module. The algorithm for sending AFECG data file and thefetal heart bit sound file are almost same except, the data types
employed in each cases are slightly different.

Fig. 5 (c) shows the flowchart for chat session module. The
logical flow of this module is almost same as the file send andfile receive modules. The significant difference is that there isa write to and read from I/O stream mechanism are used in thecase of chat module.

Fig. 5(a) and Fig. 5(b) are the flowchart for send file and filereceive modules respectively. Based on the Figure 5a, the firststep is to open a socket for connection. A socket is one
endpoint of a two-way communication link between two
programs running on the network [4]. It is bound to a port
number so that the TCP layer can identify which data is send
to which destination. Then, an I/O stream is opened forconnection. The socket I/O stream will remain open until the
LPMS is exited.

3. FECG EXTRACTION AND DETECTION
In this article, our intention is to focus on the network part ofthe systems then, we feel to need a brief description on theothers parts of the systems like, FECG Extraction anddetection from the AFECG signal. The algorithm of thisportion is a crucial part of the ambulatory recorder forprocessing the AFECG signal. For measuring the FHR andmaternal heart rate (MHR) [5] its needs to extract and detect
the fetal and maternal R peaks from the AFECG signal. The
algorithm has been developed for this purpose based on thedigital filtering, adaptive threshold, and statistical properties of
the time domain signal as shown in Figure 6. Equation 1
expresses an AFECG signal obtains from the maternal
abdomen wall.
A(t )  R (t ) M (t )  F (t )  N (t )  B (t )(1)

Where, R(t) is modulation functions from respiration, M(t) is
MECG, F(t) is FECG, N(t) is noise and B(t) is low frequency
baselines (< 10 Hz) from the muscle movement. After filtering
the AFECG signal, the equation (1) can be written as equation(2). This equation consist only the maternal and fetal ECG
signals with a small noise.
A (t )  M (t )  F (t )  N (t )

(a)

(b)

(2)

The algorithm then detects the maternal QRS complexes bythreshold the maximum value of the match filter output [6]. A
maternal ECG template is formed from the successfully
detected the maternal QRS complexes using the R peak as thefiducial point [7]. This routine first measures the slope of thematched filter output by the equation (3).

(c)

Figure 5. (a) Send file module flowchart (b) Receive file module flowchartand (c) Chat module flowchart.

y ' ( n )  y ( n )  y ( n  1)

The algorithm for receiving the file is almost similar with the(3)

A maxima in the sample (n-1) is considered when the slope575

changes from y’(n-1)  0 to y’(n)  0. This duration isassumed minimum fetal QRS duration. In this search routine,
the time instant of the successive three maxima such as max1,
max2, and max3 are stored within an R wave search interval.
When the value of max1 is greater than of the threshold, an R
peak is assumed to be found and the value of max2 isconsidered as noise peak. If the value of max2 is validated as
the R peak then the value of max3 is kept as a noise peak. In
this way the threshold is continually updated the R peak andnoise levels. A moving average routine is used to average the
MECG template, the fetal QRS template, noise peaks and the
RR intervals.

amplitude of FECG signal within the AFECG signal. If the
FECG is very small (less than 3 V) accurate detection of the
fetal R peaks would not be possible.
The TCP/IP provides a reliable and point-to-point
communication channel for the client-server application fortransfer the AFECG signal through the Internet.

4. RESULTS AND DISCUSSION
A number of cases examined by transmission and reception of
the data files. The received data after processed comparedwith the sent data and found almost same. Fig. 6 shows one of
the result indicate the heart rate of a patient at 38 weeks ofpregnancy. In this case, the AFECG signal is processed at the
remote terminal by our developed program. It compared withthe practical data and found the MHR and FHR recording
performance is almost same. In addition, the chatting sectionof the network program also worked well. The developed
algorithms can extract both the maternal and fetal heart rates
in real-time utilizing a single-lead configuration. In thiscomparison, we used a simultaneous ECG signals from thesame subjects. A few snap shot has been representing here in
Fig. 7 and found they are almost same nature of graphs.

Figure 7. Samples comparison- above three figures are our CARDIC based
ECG records and below three figures are the commercial BIOPAC System
ECG records

6. REFERENCES[1] M. A. Hasan, M. B. I. Reaz, M. I. Ibrahimy, M. S. Hussain, J. Uddin.
Detection and Processing Techniques of FECG Signal for Fetal
Monitoring”, Biological Procedures Online, 2009, pages 33.
[2] R. Vaze, R. W. Heath Jr, On the Capacity and Diversity-Multiplexing
Tradeoff of the Two-Way Relay Channel, computer Science, Information
technology, 2008, pp. 1-36.
[3] C. J. Konig, M. Buhner, G. Murling, Working Memory, Fluid Intelligence,and Attention Are Predictorsof Multitasking Performance, but
Polychronicity and Extraversion Are Not, Human Performance, 2005, vol.
18 (3), pp. 243-266.
[4] G. J. Fakas, A. V. Nguyen, D. Gillet. A Collaborative and Cooperative
Learning Environment for Web-based Experimentation, The electroniclaboratory journal, 2005, vol. 14 (3), pp. 189-216.
[5] R. W. F. Campbell, P. Gardiner, A. Murray. Automatic Analysis of
Ambulatory ECG Records, European Heart Journal, 1984, vol. 5, pp. 3134.
[6] A. Matonia, J. Jezewski, T. Kupka, K. Horoba, J. Wrobel, A. Gacek. The
Influence of Coincidence of Fetal and Maternal QRS Complexes on Fetal
Heart Rate Reliability, Medical and Biological Engineering andComputing, 2006, pp. 393-403.

Figure 6. FHR and MHR traces from patient (week 38)[7] F. Ahmed, M. I. Ibrahimy, M. A. M. Ali, E. Zahedi. A Portable Recorderfor Long-Term Fetal Heart Rate Monitoring, Microprocessors andMicrosystems, 2002, vol. 26(7), pp. 325-330.

5. CONCLUSION
The performance of the system for determining the FHR
depends upon the FECG signal, which is obtained bysubtracting the average MECG from the AFECG signal. It isfound that the performance of the system also depends on the576

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Effects of Gypsum on the Swelling and Dispersion Behavior ofFine Grained Soils from Sudan
Abdul Karim M. Zein and Ahmed M. Elsharief,
Building and Road Research Institute (BRRI), University of Khartoum, SUDAN
worldwide on the characterization of dispersive behavior ofsoils and evaluation of the methods and techniques which
may be adopted for their treatment. The experience withdispersive soils when used for construction of civilengineering works is rather short in Sudan and there are only
few reported studies and investigations [5] in this field.

ABSTRACT: This paper presents a study on the effect of gypsum
addition in improving the swelling and dispersion characteristics offour compacted expansive and dispersive soil samples from Sudan.
Gypsum was added to the soil samples studied in amounts varied
from 2 to 9% after being prepared at specified placement conditions.
Swelling tests performed on two highly expansive soils included thefree swell, linear shrinkage, swell percent and swelling pressure.
The double hydrometer, pinhole and chemical tests were used forthe evaluation of soil dispersion behavior in two samples of theRiver Nile alluvial deposits. The study results showed that theswelling potential of the highly expansive clay soils could be
reduced significantly upon their treatment with around 5 to 6% of
gypsum admixture but adverse effects may result if this gypsum
content is exceeded. The results of dispersion tests indicated that theaddition of 2 to 4% of gypsum resulted in an improvement in thedispersive characteristics such that the dispersion degree in the twosoils changed from dispersive or highly dispersive to intermediate
or completely non-dispersive states.
Keywords: Expansive soils; dispersive soils; gypsum; soil treatment;swelling potential; dispersion phenomenon

1. INTRODUCTION
Soil types that exhibit significant volume changes inassociated with changes in their moisture content are referred
to as expansive or swelling soils. Such significant changes
due to the swelling/shrinking nature are likely to cause heavy
distress or damage to engineering structures built on them.
Expansive soils are internationally widespread and of major
economic significance. The annual value of damages tovarious types of structures built on or within these soils runs
into extremely high costs worldwide. In Sudan, the expansive
soil problem has been recognized since 1967 and several
studies have been carried out to study their engineering
properties and evaluate the performance of some foundation
types placed on them [1], [2], [3]. Sudan’s climate is wholly
tropical and the expansive soils represent about 25% of thecountry area and occur within the central and eastern clay
plains. The expansive soils treatment methods currently used
include soil removal and replacement, remoulding andcompaction, pre-wetting, and chemical stabilization.
Dispersive soils may be defined as those soils that have
repulsive forces between the clay particles that exceed theattractive forces when placed in water. The tendency of theclay soils to disperse or de-flocculate depends upon themineralogy, soil chemistry, the salts dissolved in the porewater and the water causing soil dispersion. Many earth dam
or highway embankments and hydraulic structures
constructed partially or entirely with dispersive soils havesuffered serious erosion problems or have failed as reported
in published literature [4]. There are many studies carried out

Most of the failures and distresses that have occurred inengineering structures associated with expansive anddispersive problems are related to the presence of the clay
mineral montmorillonite in such soils. The mechanisms ofsoil swelling and soil dispersion are dependent on the
thickness of the diffusion double layer and the concentrationand valence of absorbed ions around clay particles. Soilswelling occurs due to the unbalanced electrical charge andcation exchange capacity produced by sodium based
montmorillonite. However, soil dispersion seems to be a
result of the presence of sodium ions adhering tomontmorillonite particles rather than the sodium present in
the pore water [6]. Soil tendencies to swell or disperse can be
reduced by replacing sodium ions by compounds which may
produce cations of lower ion exchange capacity to form amore balanced electrical charge in soil structure.
Replacement of monovalent sodium by substances withdivalent calcium ions such as lime and gypsum may lead tomarked reduction in diffuse double layer thickness which inturn leads to a lower swelling potential and reduced potential
for dispersion. By mixing lime or gypsum with a soil, thecalcium present in these substances encourages theflocculation of the soil particles thereby improving the soil
structure.
Many previous investigations were performed using chemical
admixtures to reduce the swelling potential of expansive soils
and improve the behavior of dispersive soils. However, only afew studies can be cited in published literature on the use of
gypsum stabilization for the treatment of expansive anddispersive soils.
The study presented in this paper is aiming to contribute inthis research area by investigating the effects of gypsum
addition on the behavior of selected highly expansive anddispersive Sudanese soils.
2. SOIL TYPES AND TESTING METHODLOGY
Four different soil types collected in disturbed conditions
from three different locations in the country were consideredin this study. They included two potentially expansive soils
collected from central and southern Sudan (designatedhereinafter as Soils A and B) and two dispersive soils
sampled from northern Sudan (designated as Soils C and D).
Previous studies have shown that montmorillonite is thedominant clay mineral present in all the studied soils [5], [7].

577Gypsum in a chemical form of hydrated calcium sulphate(CaSO4-2H2O) was added in predetermined quantities toportions of the soil samples and mixed properly and stored forcuring to obtain variable gypsum percentages based on the
total dry mass of the soil samples. The expansive soils
samples were mixed with 3%, 6% and 9% of gypsum,
whereas the dispersive soil samples were mixed with 2% and
4% of gypsum.
The basic index and engineering properties of the natural andgypsum treated soil samples were determined to classify theirtypes in accordance to BS1377 soil testing standards.
Compaction tests were performed according to the standard
Proctor compaction test method to determine the optimum
moisture content (OMC) and maximum dry density (MDD)
for each soil.
A concise summary is given in Tables I and Table II for thebasic properties and the compaction test results of the
expansive and dispersive soil types respectively before andafter treatment with gypsum.

A
0

3

6

9

0

3

6

9

FC (%)

83

87

88

86

80

81

80

78

CC (%)

47

14

3

2

12

10

7.5

3

LL (%)

66

58

66

65

62

58

56

60

PI (%)

36

34

41

37

38

39

37

36

OMC
(%)

27

28

28

28

23

24

24

25

MDD
(kN/m3)

14.7

14.7

14.4

14.5

15.6

15.4

15.1

14.9

C

D

Gypsum (%)

0

2

4

0

2

4

FC (%)

57

55

60

73

72

70

CC (%)

11

10

11

14

0

0

LL (%)

44

34

36

37

33

36

PI (%)

18

7

7

11

7

7

OMC (%)

21

21

21

21

22

22

MDD (kN/m3)

16.1

16.7

16.8

16.2

16.5

16.4

Four direct and indirect tests were performed to evaluate theswelling potential of the two compacted expansive soils intheir natural and gypsum treated states namely; linear
shrinkage, free swell, swell percent (one dimensional volume
change) and swelling pressure tests. The former two tests areused for the identification and classification of the swelling
potential of expansive soils. The swell percent and swellingpressure tests were performed on samples prepared in thelaboratory at two different initial placement conditions; one at
OMC (denoted as Samples A1 and B1) and the other at a dryof optimum moisture content but with the dry density
approximately maintained at the MDD value of each soil
(denoted here as Samples A2 and B2). The swell percent andswelling pressure tests were performed in an oedometer
apparatus on soil specimens, 76mm in diameter and 20mm
thick. During testing, the samples were allowed to saturate
under the conditions of free vertical swelling under a nominal
load (0.125kN) in the swell percent test and a full restrained
condition in the constant volume swelling pressure test.

B

Gypsum(%)

Soil type

3.2 Soil Conditions and Swelling Properties Considered

Table I Index and compaction properties of expansive soils
Soil type

Table II Index and compaction properties of dispersive soils

The main aspects relating to the effects of the gypsum on theswelling characteristics and dispersive behavior of study soils
are described and discussed separately in the following
sections.
3. EFFECT OF GYPSUM ADDITION ON SWELLINGBEHAVIOR OF EXPANSIVE SOILS
3.1 General
The behavior of expansive soils is normally evaluated interms of two main swelling potential components; volume
change and development of swelling pressure due to anincrease in their moisture content. These two parameters are
determined in the laboratory using an oedometer apparatus
and the test results are used for the prediction of the soil heave
under field conditions.

3.3 Analysis and Discussion of Swelling Test Results
To evaluate the effects of gypsum addition on the swellingcharacteristics of Soils A and B, the various classification andswelling test results were plotted against gypsum content asshown in Figures 1 through 4. The variations of the plasticity
index (PI), linear shrinkage (LS), free swell (FS) and clay
content (CC) fraction with added gypsum percents are asshown in Fig. 1(a) and 1(b) for Soils A and B respectively.
As may be noted from the graphical relationships shown in
Fig. 1, the addition of up to 9% gypsum indicated small
reductions in the values of the plasticity index (PI) for bothsoil types. However, mixing gypsum with soil samples has ingeneral produced more significant changes in the values ofthe other three swelling indicative indices. The addition of 3
to 9% gypsum content resulted in reductions in the free swell(FS) by 4 to 25%, linear shrinkage (LS) by 4.3 to 50% andclay content (CC) by 16.7 to 96%. It may also be noted thatthe effect of gypsum addition on the swelling potential578

indices is more evident in the samples of Soil B than in thoseof Soil A.

Table III Swelling potential classification of Soils A and B
before and after treatment with gypsum
Swelling
potential
rating method
Holtz and
Gibbs [8]Altmeyer[9]
Chen [10]Soil
Index orproperty

Soil A+ % gypsum

Soil B+ % gypsum

0

3

6

9

0

3

6

9

Free Swell

H

H

H

H

H

M

M

M

LinearShrinkage

H

H

H

H

H

H

M

M

PI

H

H

H

H

H

H

H

H

L

L

L

L

L

H

M

L

L

Van
PI andVH M L
L M
derMerwe[11] CC
ThomasSwell
H
M M M Het al [12]percent
ThomasSwelling
M
M M L Met al [12]pressure
Swelling potential ratings: VH =Very high, H = High,M = Medium/ moderate/marginal and L = Low

Figures 2 and 3 show the variations in the swell percent andswelling pressure values respectively with added gypsum forthe four samples (two from each soil) initially prepared at
different placement conditions.

Figure1: Swelling indices versus added gypsum content for
Soils A and B.
To investigate the effects of gypsum addition on soil swellingin a rational manner, six published classification schemes
were used to evaluate the swelling potential on the basis of themeasured properties before and after treatment. A summaryof the soil swelling potential rating comparisons for thenatural and treated soil samples made according to theselected methods is given in Table III.
It may be noted from Table III that similar swelling potential
ratings were revealed by the three classification methods
based on single indices (PI, FS and LS) in Soil A even for
those samples mixed with 9% gypsum but the effects ofgypsum in reducing the swelling potential of Soil B are
obvious. The schemes based on soil indices combinations (e.g.
PI and CC) and measured swelling parameters indicate
clearly that the swelling potential in both soil types changed
from a high or very high degrees to low or medium at 3%
gypsum content and to mostly low swelling potential at 6%
and 9% percentages. Based on these observations, it may bedisclosed that the degree of swelling potential of highly
expansive soils can be significantly reduced through mixing
with 3 to 6% of gypsum.

Figure 2 Swell percent versus gypsum content
The trends of the swell percent versus gypsum percent
depicted in Figures 2 indicate significant reductions inswelling potential in three samples of both soil types upontreatment with 5 to 6% gypsum. In Sample A2 of Soil A, onlya slight reduction in the swell percent was noted for gypsum
content ranging from 0 to 3%. The observed swell
percent-gypsum content trends suggest that there is optimum
5-6% gypsum content which seems to effectively reduce theswelling potential of highly compacted expansive soils placedaround their OMC and MDD conditions. However, gypsum
contents exceeding 6% produced an adverse effect on
swelling potential particularly in the two samples compacted
at OMC whereby the swell percent increases considerablywith increasing gypsum content.

579

4. EFFECT OF GYPSUM ADDITION ON THE
BEHAVIOR OF DISPERSIVE SOILS
4.1 General
Dispersive soils cannot be identified through visualinspection or by using the results of standard laboratory soil
classification tests and therefore some other special tests havebeen developed for this purpose. These include; the doublehydrometer test, the pinhole test, the crumb test and thesodium absorption ratio (SAR) based tests.

Figure 3 Swelling pressure versus gypsum content
Somewhat different swelling pressure versus gypsum content
trends were noted in Figure 3 for the soil samples compacted
at OMC and those compacted dry of OMC in both soil types.
In the two samples compacted at OMC, the swelling pressure
remained unchanged up to 3% and 6% of gypsum content inSoils A and B respectively but then it dropped to 50% of themagnitudes corresponding to untreated soil samples.
Reductions in the swelling pressure values exceeding 60%
have been reported in another study [13] to occur as a resultof adding 6% gypsum to expansive soil samples from India
compacted at various dry densities.
In the samples compacted dry of OMC, there was no change
in the swelling pressure values up to 3% of gypsum content
but beyond this the swelling pressure increased steadily to 6%
and 9% gypsum content in Soils A and B respectively.
In other words, the effect of gypsum content on the swellingpressure is generally conforming to that noted in the case ofswell percent of the two samples compacted at OMC and
MDD in both soil types. Fortunately, most of the design codesand specifications applied in practice call for the placement ofcompacted soil fill at field placement conditions equal orclosely comparable to the OMC and MDD for mostengineering projects.
Based on the finding of this study, gypsum addition proved to
be an effective technique in reducing the swelling potential ofhighly expansive soils placed at such placement conditions.
However, gypsum addition does not seem to be an effective
measure for treatment of expansive soil samples compacted atmoisture contents below OMC values. Perhaps this may be
attributed to the complex nature of soil swelling phenomenon.
The behavior of expansive soils can best be evaluated byconsidering all the physical, chemical, and mineralogicalfactors. The role and importance of each of these factors incontributing to the total swelling magnitudes measured islargely dependent on the soil type and initial soil placement
conditions. Further investigations are required to study theeffect of placement conditions on the swelling behavior ofcompacted expansive soils that can be achieved by mixing
with gypsum.

4.2 Materials and Sample Preparation
Separate portions of the two samples representing Soils C and
D were mixed with 2% and 4 % of gypsum and subsequently
used for performing the required dispersion tests. The
dispersion behavior of Soils C and D was evaluated by threedifferent methods namely; the double hydrometer test,
pinhole test described in Clauses 6.4 and 6.2 in Part 5 of theBS1377 respectively and chemical analysis done following
the methods proposed by Sherard et al. [14] shown in Fig. 4.

Figure 4 Relation between dispersion and soil pore waterchemistry [14]Figure 5 shows the double hydrometer test results for the
samples of the CL type material (Soil C) treated with 2 and
4% gypsum tested with and without dispersant agents.
Treated samples of Soil D were not tested by the doublehydrometer method because of their extremely low clay
fractions.
4.3 Analysis and Discussion of Results
4.3.1 Effects of Gypsum on the Soil Gradation and Index
Properties
Table IV shows the effects of mixing Soils C and D with 2
and 4% gypsum on the fines content (FC), clay content (CC),liquid limit (LL) and plasticity index (PI). For Soil C samples
mixed with gypsum, there was at most a slight increase in the
FC whereas the CC did not change compared to the values
corresponding to the untreated samples implying that theaddition of gypsum did not seem to have caused any
aggregation in this material. The LL and PI of samples treatedwith 2% gypsum were noted to be much lower than those of
the untreated Soil C sample. Increasing the gypsum content

580

from 2 to 4 % did not lead to any further changes in the
Atterberg limits test results.

Figure 5 Double hydrometer test results for Soil C
For Soil D, the FC of treated sample indicated no change after
gypsum treatment but the CC decreased significantly from
14 % to zero. Practically the same LL values were obtained
for untreated and treated samples of Soil D however there wasa small reduction in the PI value compared to its value in theuntreated material.
4.3.2. Effect of Gypsum Treatment on Soil Dispersion
A summary of the results of dispersion evaluation for the
samples of Soils C and D in their natural state and after being
treated with 2 and 4% of gypsum is given in Table IV.
Table IV: Summary of soil classification and dispersion
evaluation based on various test results.
Soil typeC

D

Gypsum
(%)

0

2

4

0

2

4

FC (%)

57

55

60

73

72

70

CC (%)

11

10

11

14

0

0

LL (%)

44

34

36

37

33

36

PI (%)

18

7

7

11

7

7

DH

D

ID

ID

ID

NA

NA

Pinhole

D1

ND2

ND1

D2

ND1

ND1

Chemical

D

D

ND

D

ID

ND

D1 = Highly dispersive, D2 = Dispersive ID = Intermediate dispersion,
ND2 = Non-dispersive, ND1 = Completely erosion resistantNA = Not appropriate, DH = Double hydrometter

The double hydrometer test results showed that the treatedsamples of Soil C has an intermediate degree of dispersion
whereas the pinhole test results revealed that the treatedsamples were non-dispersive (ND2 dispersion category). On
the other hand, the chemical analysis of the pore water salts inthe material revealed that the material was still dispersive
even after being treated with gypsum.
The pinhole test results showed that the treated samples ofSoil D were completely erosion resistant (ND1 category)material. On the other hand, the chemical analysis of the porewater salts in the treated material revealed that the degree ofdispersion improved upon treatment with gypsum fromdispersive in its natural state to an intermediate degree ofdispersion for 2% gypsum and almost non-dispersive for 4%
gypsum. The evaluation based on chemical analysis results is
relatively consistent with that based on the pinhole test results.
The double hydrometer test was considered inappropriate forassessing the dispersion behavior in samples of Soil D treatedwith gypsum.
5. CONCLUSIONS
On the basis of the results of the study undertaken on theeffects of added gypsum on the geotechnical characteristics
of two highly expansive and two potentially dispersivematerials from Sudan presented in this paper the following
conclusions are drawn.
a) Expansive Soilsi) Gypsum addition has produced noticeable changes in theswelling potential of the two highly expansive soils (soils A
and B). The swelling potential ratings made according tocertain widely used published classification schemes changed
from high or very high in the natural states to a low or
medium after treatment with 3 to 9% of gypsum.
ii) The swell percent of the untreated samples of Soils A
and B compacted at OMC and MDD decreased considerably
to about 50% and 10% respectively upon mixing with 5-6%
gypsum. This gypsum content appears to represent an
optimum value which may be used for the stabilization ofhighly expansive soils. However, gypsum contents exceeding
6% caused an increase in measured swell percent values inboth expansive soil types, thus producing an adverse effect on
their swelling behavior.
iii) The results of gypsum addition on the swelling pressureof the two soils were consistent with those noted in the case ofswell percent of the samples compacted at OMC and MDD.
However, gypsum addition does not seem to be an effective in
reducing the swelling pressure of expansive soil samples
compacted dry of the OMC values.
b) Dispersive Soils
i) The results obtained from the various dispersion tests
performed (i.e. the double hydrometer, pinhole and chemical
tests) clearly indicated that Soils C and D soil samples arepotentially dispersive in their natural state with a degree ofdispersion evaluated to vary from dispersive to highly
dispersive.
ii) Additions of 2 and 4 % of gypsum did not affect thegradation or the clay content of Soil C, suggesting that the
samples did not experience aggregation formation. As forSoil D, the addition of gypsum significantly reduced the clay
fraction. The addition of gypsum considerably reduced the581

plasticity of Soil C, but the reduction in the plasticity of Soil
D was less significant.
iii) The results from the three different dispersion tests
performed showed that, for both soil types studied, theaddition of gypsum resulted in significant improvements intheir degree of dispersion such that the treated samples
became non-dispersive.
6. REFERENCES[1] El-Jack, S. A. (1967), Physico-chemical mechanism of clay swelling andidentification of swelling potential, Journal of Sudan Engineering Society No.
12[2] Osman, M. A. and Charlie, W. A. (1984), Engineering properties of
expansive soils in Sudan, Proc. 5th Int. Conf. on Expansive Soils, Adelaide,
pp. 311-315.
[3] Osman, M. A. and Hamadto, M. E. (1984), Damage assessment ofresidential houses built on expansive soil areas in Sudan. Proc. 5th Int. Conf.
on Expansive Soils, Adelaide, pp. 222-226.
[4]Paige-Green, P. (2008), Dealing with road subgrade problems in South
Africa, Proc. 12th Int. IACMAG Conf. Geo, India pp 4345-4353[5] Zein, A.K.M. and El Sharief, A.M. (2011), Effects of lime and Gypsum
additives on the dispersive behavior of two River Nile deposits. Sudan
Engineering Society Journal, Vol. 57, No. 1, pp 29-36.
[6] Indraranta, B. (1991), Erosion of dispersive soil in North East Thailand,
Geotechnical Engineering, Vol. 22, pp 1-3.
[7] Ahmed M. M. (2008), Swelling potential of some expansive soils from
Sudan, M.Sc. Thesis, Faculty of Engineering, University of Khartoum.
[8] Holtz, W.G. and Gibbs, H.J., (1956), Engineering properties of expansive
clays. Transactions ASCE, 121:641-677.
[9] Altmeyer, W.T. (1955), Discussion of engineering properties of
expansive clays. Proc. ASCE, 81(Separate No. 658):17-19.
[10] Chen, F.H. (1988), Foundation on Expansive Soils, Elsevier Science
Publishers, N.Y.
[11]Van der Merwe, D.H., (1964), The prediction of heave from theplasticity index and the percentage clay fraction of soils. Civil Engineering,South Africa 6:103-107.
[12] Thomas, P.J., Baker, J.C. & Zelazny, L.W. (2000) An expansive soil
index forpredicting shrink-swell potential. Soil Sci. Soc. Am. J. 64, 268-274.
[13] Ameta, N.K., Purohit, D.G.M. and Wayal, A.S. (2007), Characteristics,
problems and remedies of expansive soils of Rajasthan, India. 2007EJGE.
[14] Sherard, J.I. Dunnigan, L.P.and Decher, R.S. (1977), Some engineering
problems with dispersive soil, ASTM STP 623: 3-12.

582

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Comparison of Bearing Capacity Equations for Vertical Central Loading
Muhammad Awais Shafique and Tanvir Iqbal Qayyum, Department of Transportation Engineering andManagement, University of Engineering and Technology, Lahore, Pakistan
ABSTRACT: Bearing Capacity is the measure of theload that the soil can sustain when a structure is built
over it. Building codes, field tests, laboratory tests andanalytical equations are the four main sources fromwhich the bearing capacity of soils can be taken.
Researchers have come up with a number of analyticalequations over the years for the determination of bearing
capacity. This study addresses the bearing capacity
equations for vertical central loading ranging from
Rankine (1885) to Padmini (2008). The data of 13 load
tests done by various researchers were used to evaluate
the bearing capacity theories studied. The bearing
capacity values calculated from the equations werecompared with the test values. Moreover mean squared
error was also determined. The evaluation shows that inmost cases Meyerhof’s Bearing Capacity equation
provides much closer results to the actual ones, andtherefore should be preferred except for square footings
where the equation by Balla yields better results.

1.

calculated using different equations for the primarydata provided by the researchers and thencompared with the actual bearing capacities
mentioned. The mean squared error for eachapproach was calculated and the most suitable
theory was then figured out.
3.

BEARING CAPACITY EQUATIONS

Rankine was perhaps the first researcher who
proposed an equation for calculation of the bearing
capacity of cohesionless soils. He derived his
equation from the Rankine’s earth pressure theory
[4].
(1)
Where

INTRODUCTION= Ultimate bearing capacity= Density of soil
= Foundation depth= Angle of internal friction

Bearing capacity of soil is the capacity of soil tobear the load applied by the structures constructed
over it. It is perhaps the most important aspect ofsoil.

Rankine was unable to accommodate the effect offooting size on bearing capacity. Pauker came up
with a similar equation for sands as the one given
by Rankine. His equation became very popular but
had the same limitations as that of Rankine’s [5].

Bearing capacity values are obtained by four main
sources; building codes, field tests, laboratory tests
and analytical methods. The use of analyticalequations for computing bearing capacity is one of
the widely used methods. Over the ages a lot of
researchers have developed different concepts
about the bearing capacity calculations and hence
have added to the wide range of equations which
can be used for bearing capacity calculations. The
problem with so many theories is to decide which
may give the best result for a particular soil, under
given conditions.

45

The Pauker-Rankine equation was modified by
Bell to be used for cohesive-frictional soil. Again
the effect of footing size was neglected [6].
2

The scope of work covers the bearing capacity
equations dating back from Rankine’s theory to thepresent era bearing capacity equations. The aim of
the study is to evaluate various equations undervertical central loading and propose the mostsuitable.
2.

(2)

1

(3)

Where
45

(4)

The first prominent achievement in the field of
bearing capacity calculation was by Terzaghi. He
studied failure zones beneath strip footings andthereby developed his famous bearing capacity
theory. The equation proposed by Terzaghi
incorporated the influence of footing size on
bearing capacity in the form of shape factors [7].

METHODOLOGY

Various equations developed over the years for the
calculation of bearing capacity under verticalcentral loading were studied. The data of 13 load
tests carried out and documented by Muhs andWeiss [1], Melovic [2] and Muhs et al [3] was used
to evaluate and compare the bearing capacity
equations studied. The bearing capacities were

1

583

2

(5)

Wherecq
B
Nc, Nq, Nγ
S c, S γ

Balla developed graphs (Fig. 1, 2 and 3) for
calculating the bearing capacity factors to be usedin Terzaghi’s Equation [10].

= Cohesion= Overburden soil pressure= Width of footing= Bearing capacity factors= Shape factors

Hu studied the base angle of the triangular wedge
below a rough foundation and proposed his own
graph (Fig. 4) for determination of bearing capacity
factors [11].

Meyerhof modified the equation given by Terzaghi
by adding depth factors and inclination factors. So
the equation became equally applicable for inclined
loads. Furthermore the values of bearing capacity
factors were different from those given by Terzaghi[8], [9].
qult = cNcScdc + qNqSqdq + 0.5γBNγSγdγ

Terzaghi’s equation was simplified by Krizek, whodeveloped an easy equation valid for = 0 to 35o[12].
qult = [(228+4.3  )c+(40+5  )q+3  γB]/(40-  ) (8)

(6)

Where
d c , d q, d γ= Depth factors

Fig. 1:

Balla’s bearing capacity factor Nc

Fig. 3:

Balla’s bearing capacity factor Nγ

Fig. 2:

Balla’s bearing capacity factor Nq

Fig. 4:

Hu’s bearing capacity factors

584

Meyerhof’s equation was further extended byHansen who included base factors and groundfactors in addition to some slight changes [13].
0.5Based on soil strength parameters, Bolton and Lau
established a bearing capacity calculation
and , given in Table
procedure. The values ofI, were proposed for circular and strip footings
subjected to vertical loading [15]. Qayyum
experimented on sand under rapid loading andproposed a general ultimate bearing capacity
equation applicable for vertical as well as inclined
central and eccentric loading [16].

(9)

Where
,
,

,
,

= Base factors= Ground factors

The simplified form of the equation for verticalcentral loading is,

Vesic proposed a similar equation like the one
suggested by Hansen. Both these equations areapplicable for vertical as well as for inclined
loading [14].

0.45

Salgado et al (2004) determined the bearing
capacities of square, rectangular, strip and circular
foundations in clay accurately based on finite
element limit analysis. Using these analyses precise
values of shape and depth factors were proposed[17].

Nc and Nq by Bolton and Lau

Table I:

(10)

Nγ

Nq
Smooth or
rough
Strip Circle

Strip

CircleStrip

Circle

5

1.57

1.65

0.09

0.06

0.62

0.68

10

2.47

2.80

0.29

0.21

1.71

1.37

15

3.94

4.70

0.71

0.60

3.17

2.83

20
25

6.40
10.7

8.30
15.2

1.60
3.5

1.30
3.0

5.97
11.6

6.04
13.5

30

18.4

29.5

7.7

7.1

23.6

31.9

C1 and C2 are determined from Table II

31

20.6

34.0

9.1

8.6

27.4

38.3

32

23.2

39.0

10.7

10.3

31.8

46.1

33

26.1

45.0

12.7

12.4

37.1

55.7

Padmini et al developed neurofuzzy models for theprediction of ultimate bearing capacity of shallow
foundations on cohesionless soils [18].

34
35

29.4
33.3

52.2
61.0

15.0
17.8

15.2
18.2

43.5
51.0

67.6
82.4

Table II:

36

38

71

21

22

60

101

37

43

83

25

27

71

124

38

49

99

30

33

85

153

39

56

116

36

40

101

40
41

64
74

140
166

44
53

51
62

42

85

200

65

43

99

241

44

115

45

(deg)

Smooth

Rough

1

0.27

(11)

1

(12)

Where

Regression constants in (12) for the
shape factor

B/L

C1

C2

190

1 (circle)

0.163

0.210

121
145

238
299

1 (square)

0.125

0.219

0.50

0.156

0.173

78

176

379

79

99

214

480

295

97

125

262

619

135

359

120

160

324

803

0.33
0.25
0.20

0.159
0.172
0.190

0.137
0.110
0.090

46
47

159
187

444
550

150
188

210
272

402
505

1052
1384

48

222

686

237

353

638

1847

49

265

864

302

476

815

2491

50

319

1103

389

621

1052

3403

51

386

1427

505

876

1373

4710

52

470

1854

663

1207

1812

6628

585

4.

DATA COLLECTION

Data from 13 different Test scenarios was collectedfrom literature. The researchers have actually
performed these load tests on different soils anddetermined the bearing capacities. The data setstaken from [1], [2] and [3] are given in Table III,IV and V respectively.

Table III: Data from Muhs and Weiss (1969)
Data Set

1

2

3

4

Depth, Df(m)

0

0.5

0.5

0.5

Width, B (m)

0.5

0.5

0.5

1

Length, L (m)

2

2

2

1

Unit wt. γ (KN/m3)

15.69

16.38

17.06

17.06

Angle of internal friction, φ

38.5

36.25

40.75

38.5

6.37

3.92

7.8

7.8

1059.121196.41

2373.213236.20

Cohesion, c (KPa)
2

Actual Bearing Capacity, qu (KN/m )Table IV: Data from Melovic (1965)
Data Set

5

6

7

8

Depth, Df(m)

0.4

0.5

0

0.3

Width, B (m)

0.71

0.71

0.71

0.71

0.71

0.71

0.71

0.71

Unit wt. γ (KN/m )

17.65

17.65

17.06

17.06

Angle of internal friction, φ

22

25

20

20

12.75

14.7

9.8

9.8

402.07539.37215.75254.97

Length, L (m)3

Cohesion, c (KPa)
2

Actual Bearing Capacity, qu (KN/m )Table V: Data from Muhs et al (1969)
Data Set

9

10

11

12

13

Depth, Df(m)

0.3

0

0.3

0

0.3

Width, B (m)

0.6

0.6

0.6

0.6

0.6

1.2

1.2

1.2

1.2

1.2

Unit wt. γ (KN/m )

9.85

10.2

10.2

10.85

10.85

Angle of internal friction, φ

34.9

37.7

37.7

44.8

44.8

Cohesion, c (KPa)

0

0

0

0

0

Actual Bearing Capacity, qu (KN/m2)

270

200

570

860

1760

Length, L (m)35.

cases. The safe values or under-estimated values
closest to the actual ones were figured out for eachcase. The mean squared error was then calculated(Table VII). Although the mean squared error incase of cohesionless soil is minimum for Qayyum’s
equation but as it is only applicable for surface
loading so it is therefore neglected.

EVALUATION AND COMPARISON

Using the data collected from various researchers,
the bearing capacity values were calculated using
the equations studied. The comparison is given in
Table VI. It can be seen from the table that noindividual equation gave the best results in all the586Table VI: Comparison of Bearing Capacity Equations

Primary
data

Bearing
Capacityequation

Data Set

1

2

3

4

5

6

7

8

9

10

11

12

13

Depth, Df(m)

0

0.5

0.5

0.5

0.4

0.5

0

0.3

0.3

0

0.3

0

0.3

Width, B (m)

0.5

0.5

0.5

1

0.71

0.71

0.71

0.71

0.6

0.6

0.6

0.6

0.6

Length, L (m)

2

2

2

1

0.71

0.71

0.71

0.71

1.2

1.2

1.2

1.2

1.2

Unit wt. γ (KN/m3)

15.69

16.38

17.06

17.06

17.65

17.65

17.06

17.06

9.85

10.2

10.2

10.85

10.85

Angle of internal friction, φ

38.5

36.25

40.75

38.5

22

25

20

20

34.9

37.7

37.7

44.8

44.8

Cohesion, c (KPa)

6.37

3.92

7.8

7.8

12.75

14.7

9.8

9.8

0

0

0

0

0

Actual Bearing Capacity, qu (KN/m2)

1059.11196.42373.23236.2

402.1

539.4

215.7

255.0

270

200

570

860

1760

Terzaghi

883.3

896.5

2189.7

1985.0

426.4

634.3

243.0

281.1

239.8

205.5

386.7

915.7

1463.1

Meyerhof

807.1

1009.72597.82789.2

468.3

744.1

225.5

293.3

266.3

224.7

439.5

1058.91801.4

Balla1009.8889.0

2157.5

1941.8

355.2

533.9

195.4

233.5

253.0

228.3

409.6

1017.51564.9

Hu

904.4

859.7

1932.1

1887.6

338.8

476.2

199.2

230.9

298.5

336.6

465.1

992.8

1367.1

Krizek

-

-

-

-

333.3

488.5

190.2

226.0

245.6

-

-

-

-

Hansen

711.8

949.2

2293.82218.1

490.8

779.8

218.8

303.7

220.0

130.9

341.2

504.1

1130.6

Vesic

795.3

1025.52462.02425.2

504.0

800.3

227.6

313.4

259.8

182.1

405.4

682.5

1375.8

Bolton & Lau

777.2

782.9

1840.91746.1

333.5

472.9

181.6

214.3

245.7

247.2

391.7

1014.31440.7

Qayyum

-

-

-

-

-

-

-

-

-

205.5

-

915.7

-

Padmini et al

-

-

-

-

-

-

-

-

169

95

495

632

1986

The underlined values are safe values closest to the actual values.

587

Table VII: Mean Squared Error[3]

Muhs, H, Elmiger R, Weiß K., “Sohlreibung undGrenztragfähigkeit unter lotrecht und schräg belasteten
Einzelfundamenten”, Deutsche Forschungsgesellschaft
für Bodenmechanik (DEGEBO), Berlin. HEFT 62, 1969.

Mean Squared error

Bearing
Capacity
Equationsc-φ soil(1 - 8)φ soil(9 ‐ 13)Overall[4]

Rankine, W. J. M., “A manual of applied mechanics”,
Charles Griffin and Co., London, 1885.

Terzaghi

21639425155

142840[5]

Meyerhof

49564

11789

35035

Pauker, H. E., “An explanatory report on the project ofa Sea-battery” (in Russian), Journal of the Ministry ofways and communication, St. Petersburg, 1889.

Balla

22774917942

140153[6]Hu26991540494

181676

Bell, A. L., “Lateral pressure and resistance of clay andthe supporting power of clay foundations”, Minutes,
Proceedings of the Institution of Civil Engineers,
London, vol. 199, 1915.

Krizek

-

-

-

[7]

Hansen161589116488144242

Terzaghi, K., “Theoretical Soil Mechanics”, John Wiley& Sons, New York, 1943.

Vesic

105803

41331

81006

[8]

Meyerhof, G. G., “The ultimate bearing capacity offoundations”, Geotechnique. 1951, 2: 301.

Bolton & Lau

34575632078

225111[9]

Qayyum-1568-

Meyerhof, G. G., “Some recent research on the bearing
capacity of foundations”, Canadian Geotech. 1963, J.
1(1): 16.

Padmini et al

-

25982

-

[10]

Balla, A., “Bearing capacity of foundations”, J. Soil
Mech. Found. Div., ASCE, 1962, 88(5): 13.

[11]Hu, G. G. Y., “Variable-factors theory of bearing
capacity”, J. Soil Mech. Found. Div., ASCE, 1964,
90(4): 85.

[12]

Krizek, R. J., “Approximation for Terzaghi’s bearing
capacity”, J. Soil Mech. Found. Div., ASCE, 1965,
91(2): 146.

[13]

Hansen, J. B., “A revised and extended formula forbearing capacity”,Bulletin No. 28, Danish GeotechnicalInstitute, Copenhagen, 1970.

[14]

Vesic, A. S., “Analysis of ultimate loads of shallow
foundations”, J. Soil Mech. Found. Div., ASCE, 1973,
99(1): 45.

[15]

Bolton, M. D., and C. K. Lau, “Vertical Bearing
Capacity Factors for Circular and Strip Footings on
Mohr-Coulomb Soil”, CGJ, vol. 30, no. 6, Dec, 1993,
pp. 1024-1033.

[16]

Qayyum, T. I., “Bearing capacity of reinforced andunreinforced soil under rapid loading”, Ph.D. thesis,
University of Strathclyde, Glasgow, 1995.

[17]

Salgado, R., Lyamin, A. V., Sloan, S. W. & Yu, H. S.,“Two- and three-dimensional bearing capacity offoundations in clay”,Geotechnique 54, No. 5, 2004, pp.
297–306[18]

Padmini, D., Ilamparuthi K., Sudheer K.P., “Ultimate
bearing capacity prediction of shallow foundations oncohesionless soils using neurofuzzy models”,
Computers and Geotechnics 35, 2008, pp. 33–46

After Qayyum [16] the minimum value of error isfor Meyerhof’s equation, which suggests that
Meyerhof [8], [9] gives results closest to the actual
ones. But Meyerhof [8], [9] yields over estimatedresults in case of square footings (data sets 5 – 8).
Here the equation by Balla [10] gives close andsafe results.
6.

CONCLUSION

From the above discussion
conclusions can be drawn.





7.

the

following

No single equation gives the best results
for all the cases.
Meyerhof’s bearing capacity equation
gives better results than the others in mostcases, but over estimate for squarefootings.
The equation proposed by Balla yields
good results in case of square footings.

ACKNOWLEDGEMENT

The authors are thankful to the efforts put in by
Prof. (R) Muhammad Siddique Qureshi inreviewing this work and giving valuable feedback.
8.

REFERENCES[1]

Muhs, H., and K. Weiss, “The Influence of the Load
Inclination on the Bearing Capacity of Shallow
Footings”, 7th ICSMFE, vol. 2, 1969, pp. 187-194.

[2]

MiIovic, D. M., “Comparison between the Calculated
and Experimental Values of the Ultimate Bearing
Capacity”, 6th ICSMFE, vol. 2, 1965, pp. 142-144.

588

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Pull-out Resistance of Selected Native Plants in Hong Kong
Leung, F.T.Y., Yan, W.M., Tham, L.G., Department of Civil Engineering, The University of Hong Kong, Hong
Kong, China
Hau, B.C.H., School of Biological Sciences, The University of Hong Kong, Hong Kong, China



ABSTRACT: The use of live plants to stabilize slopes is attractive tosociety due to its environmental friendliness and sustainability.
However, stability of the plant itself is a major safety concern. In thisstudy, field pull-out tests are performed on selected native woody
species that are commonly planted on slopes in Hong Kong toinvestigate their pull-out characteristics. The peak pull-out
resistance for individual species is correlated to plant height, basal
diameter, canopy size, and above-ground dry weight. It was foundthat basal diameter is a good predictor of the peak resistance. Rootarchitecture of the plants was also studied. Attempts are made to
predict the pull-out characteristics of a species from its rootarchitecture.

Keywords: Slope stability, Plants, Pull-out resistance,
Soil-root interactions
1. INTRODUCTION
Slope instability remains a major geotechnical hazards inHong Kong due to its rugged and hilly topography andfrequent rainstorms in the summer. The traditional approachof concrete-covering for slope protection cannot meet
nowadays requirement for a green and sustainable
environment. The use of live plants as a slope cover thus
becomes an attractive and favorable solution. Firstly, theplants increase the suction level of the ground soil throughwater absorption and transpiration. An increase in soil suction
enhances the shear strength of the soil and thus improves
slope stability. Secondly, the root system reinforces theground through the mechanical interaction between roots andsoil. It also protects the soil from erosion. The aim of thisstudy was to find out the pull-out characteristics of selectedwoody plant species that are commonly planted on man-madeslopes in Hong Kong.
In this study, four native woody species with high ecologicalvalues, which had good survival rates in previous
investigations in slope greening, were selected [1]. The
pull-out characteristics of the plants is studied through aseries of field pull-out tests. In addition, the height, basal
diameter, canopy size and above ground dry weight of theplant were measured and attempts were made to correlatethem to the peak pull-out resistance of the plant. Rootarchitecture was studied to shed light on the pull-out
characteristics.

2. SELECTED SPECIESTo offer a sustainable solution for slope greening, the plant
species used must be able to promote natural regeneration inorder to form a sustainable vegetation cover. Besides, plants
with high ecological values are preferable as many slopes are
situated in the countryside or even within Country Parks. As
such, native plant species is preferred. On the other hand,
many man-made slopes are quite steep (>45 degrees), large
and tall trees on slopes may become a negative attribute toslope stability. Therefore, shrubs and small trees are moresuitable for Hong Kong’s application [2]. In this study, the 4native species chosen are Rhodomyrtus tomentosa,
Melastoma sanguineum, Schefflera heptaphylla and Reevesia
thyrsoidea. The former two are shrubs and the latter two aresmall trees (Table 1). All of them have been commonlyplanted on man-made slopes in Hong Kong since the last
decade.
Table 1. The ecology of the four selected plant species
Seeddispersal
agent[5]-[7]
Bird, rat,
civet

Growthform[1]

Typicalheight[3],[4]

Fruit
type
[3]

Rhodomyrtus
tomentosa

Shrub

1-2 m

Berry

Melastomasanguineum

Shrub

1.5-3 mFleshycapsuleRat, bird

Schefflera
heptaphylla

Small
tree

5-10m

Berry

Bird

Reevesia
thyrsoidea

Small
tree

5m

CapsuleWind

Species

Ecologicalstatusin Hong Kong
[5][7][8]Commonpioneer
species on
Hong Kong’s
shrublandCommonpioneer
species on
Hong Kong’s
shrubland andsecondaryforest3. FIELD PULL-OUT TEST
3.1 Site and Equipment
The field pull-out tests were conducted in the natural sloping
terrains in Hong Kong. A tailor-made pull-out apparatus wasassembled to perform the tests. Fig. 1 shows the equipment
set-up. It comprises of a reaction frame, a 10 kN load cell
(TML TCLP-1B), a displacement transducer (Hoskin
Scientifique DP-2000C), and a winch system. The sensors are589

connected to an automatic data acquisition system (TML
TC-31K).

behavior is highly related to the root architecture of the plant,
which will be discussed more in a later section.
6

Winch system

A

Displacementtransducer

Pull-out resistance (kN)

5

Frame

Load cell

Schefflera
4

3

Reevesia
Melastoma2

Dataacquisition
system

Rhodomyrtus

B

1

0
0

50

100

150

200

250

300

350

400

Displacement (mm)Fig. 1 Field pull-out apparatus.

Fig. 2 Pull-out response of the plants.

3.2 Size Characteristics of a Plant

Regression analyses were carried out to investigate thecorrelations between the peak pull-out resistance and the size
characteristics of individual species. It is believed that withthese simple empirical correlations the peak pull-out
resistance of a plant can be quickly estimated in the field.
Table 2 summarizes the peak pull-out resistance and othersize-characterizing parameters for each species. It can be seenthat plants between 0.5-2.5 m tall with canopy size from
0.2-1.5 m are investigated. The pull-out resistance can be as
low as 0.1 kN but as large as 6 kN.

Prior to a pull-out test, some simple plant size characterizing
parameters were recorded, including height, basal diameter,
canopy size (which is defined as the average diameter of theplant crown), and above-ground dry weight. The former threeparameters can be directly measured at the site whereas thelast one can only be determined in the laboratory. The
above-ground plant parts of the samples were first cut into
pieces in the field. They were then brought back to thelaboratory for oven drying at 105 degree Celsius overnight.
The weight of the dried sample is its above-ground dryweight.

Table 2. Summary of the pull-out resistance and plant size inthis study
Species

3.3 Results and Discussions
First the above-ground part of a plant was cut. The remaining
stem was then tied to the cable in the winch system with aload cell and a displacement transducer connected. The stem
was then pull-out steadily with the displacement and pull-out
resistance recorded continuously until the whole root system
was hanged freely from the ground. A total of 10
Rhodomyrtus, 9 Melastoma, 18 Schefflera, and 10 Reevesia
are uprooted. Fig. 2 shows the typical pull-out response of theplants where local fluctuation of the pull-out resistance can be
seen. The pull-out resistance of a plant is attributed by threemajor factors: root-soil interface friction, tensile strength ofindividual root segments, and spatial distribution of the root(i.e., root architecture). When a root breakage or a root-soil
slippage occurs, a sudden drop in the pull-out resistance is
expected (see Point A in Fig. 2). Sometimes it is followed bya subsequent recuperation (see Point B in Fig. 2) due to thestrength mobilization of other root branches until the entire
root system is uprooted.
In general, Rhodomyrtus shows smaller pull-out resistancewith large displacement before the plant can be fully uprooted.
On the other hand, Schefflera and Reevesia show larger
pull-out resistance but smaller pull-out displacement (i.e.,
small displacement is required to uproot the plants). The

Peakpull-out
resistance(kN)
0.21-2.45

Height(cm)

Basaldiameter(cm)

Canopysize(cm)

Aboveground dryweight (g)58-191

0.4-1.4

18-1605-697

Melastomasanguineum

0.17-2.95

56-2100.6-4.6

19-1395-1393

Schefflera
heptaphylla

0.31-6.28

55-2500.9-7.1

25-1279-759

Reevesia
thyrsoidea

0.13-2.67

70-1901.0-3.029-10411-254

Rhodomyrtus
tomentosa

Fig. 3 shows the correlation between the peak resistance andindividual size parameters. In general a better correlation canbe observed between the peak pull-out resistance and thebasal diameter of the plant (see Fig. 3(b)). The peak pull-out
resistance of the plants increases as their basal diameterincreases which are more obvious in the cases of Schefflera
and Reevesia. Coefficients of determination (R2 values) are
relatively low in the correlation between the resistance andother size parameters (height: 0.17-0.86; canopy size:
0.0004-0.61; and above-ground dry weight: 0.12-0.84). Thismeans that plant height, canopy size and above-ground dry

590

weight may not be good predictors of the peak resistance.
Generally, Melastoma, Schefflera and Reevesia give larger
pull-out resistance than Rhodomyrtus (shrub). Rootarchitecture of the plant may shed light on the magnitude of
the resistance against uprooting.

Peak pull-out resistance (kN)d7
Rhodomyrtus

6Melastoma

5

Schefflera

4

Reevesia

3

4. ROOT ARCHITECTURE

2

Root architecture describes the spatial distribution of a rootsystem and it has been generally classified into three types:
plate, heart and tap as shown in Fig. 4 [9].

1
0
0

50

100

150

200

250

300

plate

Height (cm)

heart

tap

Peak pull-out resistance (kN)d7
Rhodomyrtus6

y = 0.90x - 0.68
R² = 0.75

Melastoma

5

Fig. 4. Shape of a root system: plate, heart and tap (redrawn
based on [9]).

Schefflera4

y = 1.12x - 0.86
R² = 0.88

Reevesia

3
2

(b)

stem

y = 0.64x - 0.17
R² = 0.96

y = 0.75x - 0.12
R² = 0.13

1

(a)

0
0

1

2

3

4

5

Basal diameter (cm)stem
Peak pull-out resistance (kN)d7
Rhodomyrtus

6Melastoma

5

Schefflera

4

Reevesia

(c)

(d)

stem

stem

3
2
1
0
0

50

100

150

200

Canopy size (cm)Peak pull-out resistance (kN)d7
Rhodomyrtus

6Melastoma

5

Schefflera

4

Fig. 5. Root architecture of: (a) Rhodomyrtus (top view);(b) Melastoma (top view); (c) Schefflera (side view);(d) Reevesia (side view).

Reevesia

3
2
1
0
0

500

1000

1500

Above-ground dry weight (g)Fig. 3. Correlations between the peak pull-out resistance withplant characteristics: (a) height; (b) basal diameter; (c)canopy size; (d) above-ground dry weight.

In this study, the plants were carefully dug out from theground and the root architecture was studied according to thisclassification system. Fig. 5 shows the typical root system of
the studied plants. Fig. 5(a) and (b) show a Rhodomyrtus anda Melastoma respectively which clearly demonstrate theirplate root system. Fig. 5(c) shows a Schefflera which isclassified as a heart system with many vertical, horizontal andinclined roots. Fig. 5(d) shows the root system of a Reevesia
in which a tap root system can be easily observed. Fig. 6 gives
a summary of the root architecture of the sampled plants inthis study. It shows that the tap root system dominated in591

these species except for Rhodomyrtus. In fact, all Reevesia
specimens in this investigation show the tap root system.
Previous investigations showed that plants with tap or heartroot systems give generally higher resistance against
uprooting [10]. This study confirms the findings - Schefflera
and Reevesia give noticeably higher peak resistance against
uprooting than Melastoma and Rhodomyrtus.
100%
90%

Tap, 2780%

Tap, 44[6]

Dudgeon D and Corlett RT, “The Ecology and Biodiversity of Hong
Kong,” Hong Kong: Friends of the Country Parks & Joint Publishing(HK) Company Ltd., 2004.
[7] Hau BCH and Corlett RT, “A survey of trees and shrubs on degradedhillsides in Hong Kong,” Memoirs of the Hong Kong Natural History
Society, vol. 25, 2002, pp. 83-94.
[8] Chau KC, “The Ecology of Fire in Hong Kong.” Hong Kong:
Department of Ecology and Biodiversity, The University of Hong
Kong, 2004.
[9] Stokes A, Atger C, Glyn Bengough A, Fourcaud T and Sidle RC,“Desirable plant root traits for protecting natural and engineered slopes
against landslides,” Plant Soil, vol. 324, 2009, pp. 1-30.
[10] Norris JE, Strokes A and Mickovski SB, “Slope stability and erosion
control: Ecotechnological solutions,” Dordrecht: Springer, 2008.

Root system

70%
60%

Tap, 80

Heart, 33

Tap, 100

50%
Heart, 25

40%
30%
20%

Plate, 40
Plate, 31

Heart, 12

Melastoma (16)

Schefflera (25)10%Plate, 80%Rhodomytrus (15)Reevesia (18)Species (number of test)

Fig. 6. A summary of the root architecture of the studied
species.

5. CONCLUSIONSSlope greening with native shrubs and small trees hasreceived much public attention due to its sustainability
potential. In this study, pull-out resistance and rootarchitecture of four native shrub and small tree species areinvestigated. Pull-out characteristics of individual species isdiscussed with reference to its root architecture.
6. ACKNOWLEDGMENT
The authors would like to acknowledge the support byresearch grant HKUST9/CRF/09 provided by the Research
Grants Council of the Hong Kong Special Administrative
Region (HKSAR).
7. REFERENCES
[1]

[2]

[3]
[4]

[5]

Hau BCH, So KKY, Choi KC and Chau RYH, “Using native tree andshrub species for ecological rehabilitation of man-made slopes in Hong
Kong,” in Proc. 25th Annual Seminar of the Hong Kong Institute ofEngineers, 2005, pp. 273-286.
GEO, “Technical Guidelines on Landscape Treatment andBio-engineering for Man-Made Slopes and Retaining Walls,” HongKong: Civil Engineering Department, HKSAR, 2000.
Hu QM, and Wu DL, “Flora of Hong Kong,” Vol. 2, Hong Kong:
Agriculture, Fisheries and Conservation Department, HKSAR., 2008.
ISSG, “Global Invasive Species Database,” Invasive Species Specialist
Group (ISSG) of the IUCN Species Survival Commission, National
Biological Information Infrastructure, Manaaki Whenua-Landcare
Research and the University of Auckland, 2011.
Hau BCH and So KKY, “Using native tree species to restore degradedhillsides in Hong Kong, China,” Bring Back the Forests: Policies andPractices for Degraded Lands and Forests, Sim HC, Appanah S andDurst PB, Ed. Bangkok, Thailand, FAO, 2003, pp. 179-190.

592

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Inland navigation in carrying injured and sick disaster victims in case of a
massive earthquake in the Tokyo metropolitan area
Kazuya EGAMI1), Keiichi NOTO2), Wataru MIYAZAKI2), Kazukiyo YAMAMOTO2), Takeo KONDO2)
1) ECOH CORPORATION 2) Graduate School of Science and Technology , Nihon University


ABSTRACT: The Tokyo metropolitan government plans toorganize a waterway transportation system utilizing boats and ships
to enhance transportation capacity to carry injured and sick people
during an earthquake disaster. However, the existing regional
disaster prevention plan formulated by the Tokyo government does
not stipulate specific operational procedures.
This paper investigates the prospect of the implementation ofearthquake disaster countermeasures using Inland navigation. To
begin with, we examined the presence of agreements on disaster
response procedures concluded between the municipalities in Tokyo,
who installed and administered the piers, and the organizations
representing boat and ship operators, such as fishermen’s unions,leisure fishing boat cooperative unions and marinas. We also
surveyed 252 users of Inland navigation on their awareness of
disaster prevention using Inland navigation. To conclude, we
propose a concept of an institutional design which utilizes Inland
navigation for emergency evacuation of vulnerable people affectedby an earthquake disaster.
Keywords: Inland navigation, Disaster prevention piers,
Institutional design, Regional disaster prevention plan,
Operating plan

1. INTRODUCTION
The Great Hanshin-Awaji Earthquake on 17 January 1997 led
to enhanced recognition of the importance of Inland
navigation and triggered the construction of disasterprevention piers throughout Japan. A disaster prevention pierserves as a base in reconstructing river administration
facilities and conducting relief operations following theoccurrence of flood, high tide and earthquake, which include
shipments of aid supplies, relief efforts for people in isolatedareas, transportation of disaster victims such as people unable
to get home, removal of rubbles, and transportation of relief
workers. In the Great East Japan Earthquake on 11 March
2011, however, those piers did not function well and over
100,000 people in Tokyo could not get home. It is attributed
to the fact that the regional disaster prevention plans do not
clearly stipulate the role of Inland navigation, which enhancethe disaster mitigation capacity, and the usage of disasterprevention piers. It is also caused by the lack of a plan thatconsiders vulnerable disaster victims. This study analyzes
the transportability of sick and injured people, the disasterprevention agreements among relevant parties under theregional disaster prevention plans, and people’s awareness of
disaster prevention. Finally, the institutional design on theutilization of Inland navigation and disaster prevention piers
which focuses on the vulnerable disaster victims is proposed.

2. DISASTER
PREVENTIONPIERSWITHINREGIONAL DISASTER PREVENTION PLANS
2.1 Organizations responsible for operation andadministration of disaster prevention piers
In principle, river administrators maintain disaster prevention
piers. They or park administrators are in charge of theiroperation at normal times, and so does disaster prevention
departments in time of a disaster. The regional disasterprevention plan formulated by the Tokyo government
stipulates that each ward is the main administrator of thedisaster prevention piers under the jurisdiction of Tokyo
within its region in time of a disaster, in order to realize thecomprehensive operation of the piers. The disaster prevention
piers under the jurisdiction of the national government are set
to be operated and administered by river administrators, but
the detailed are not stipulated.
2.2 Administration and operation of disaster prevention
piers in regional disaster prevention plans
As shown in Table 1, the Tokyo metropolitan government
defines in its regional disaster prevention plan that theoperation of disaster prevention piers and boats consists
mainly of (a) construction of an emergency transportation
network, (b) transportation of evacuees, (c) transportation of
medical staffs and sick/injured people, (d) transportation
alternative for victims unable to get home, and (e)transportation of materials and equipments. The specific rules
for the operation, however, are not designated in the plan. The
disaster prevention plans of the wards, the lower
administrative organs in Tokyo, specify only (a) constructionof the emergency transportation network. Each ward defines
different responses regarding the specific operation of thedisaster prevention piers and it is not clearly stipulated.
Table 1: Number of current disaster prevention piers andtheir operations

Administrative Body
Tokyo Metropolis
Chiyoda Ward
Chuo Ward
Bunkyo Ward
Taito Ward
Sumida Ward
Koto Ward
Shinagawa Ward
Ota Ward
Kita Ward
Arakawa Ward
Itabashi Ward
Adachi Ward
Katsushika Ward
Edogawa Ward593

Number of current disaster prevention piers (by administrator)

Operation of disaster prevention piers or boats

National
Prefecture Ward Private Total
government

Emergency transportation
Medical staffs /Sick and People unable to Materials andEvacuation
network
injured people
get home
Equipments

Year
2007
2007
2009
2007
2008
2009
2008
2007
2009
2008
2009
2009
2008
2008
2009

10
0
0
0
0
0
1
0
1
1
0
1
2
2
2

15
0
4
0
1
1
2
0
0
0
1
0
1
1
4

33
3
1
1
0
1
7
1
1
3
2
1
0
1
10

1
0
0
0
1
0
0
0
0
0
0
0
0
0
0

58
3
5
1
2
2
10
1
2
4
3
2
3
4
16

●
●
●
●
●
●
●

●

●

●
●

●
●

●
●
●
●
●

●

●

studying the feasibility of a water way transportation system
●
●
●
●
●
●
●

●
●

●
●
●

●
●

Table 2 shows the presence or absence of concrete
operational plans in Tokyo and its wards. Currently, Tokyo
has no uniform operational plan and has entrusted it to thewards in compliance with the article 5 of the Disaster
Countermeasure Basic Act. However each ward does not
specify the development of the concrete operational plan.
Eight among 12 wards cite it as a subject of future
investigation, and the two plan to consolidate the system in
the next fiscal year. The result of the survey also presents thatthe certain properties of a river and a bridge restrict thenavigation of the boats and it causes the inadequate
development of the plan.
Table 2: Existence of an employment planTokyo
Metropolis

Details

operators and has no adequate partnership with private sector
entities, which exposes the weakness of the transportation
infrastructure in time of an earthquake. Behind this lies thelack of radio communications networks for disasters and theinsurance system which are available in the private sector,
and the mechanism that clarifies the procedures of arranging
boats and giving instructions and orders.
Table 3: Presence or absence of disaster prevention
agreement
Municipality
Edogawa Ward
Katsushika Ward
Adachi Ward
Koto Ward
Sumida Ward
Arakawa Ward
Chuo Ward
Minato Ward
Shinagawa Ward
Ota Ward

No uniform operational plan formulated* Entrust each ward with the specific operational plans, in compliance with thearticle 5 of the Disaster Countermeasure Basic Act (obligation ofmunicipalities).
* Have no firm grasp of the navigable areas in the rivers.
No concrete operational plan at present* See it as a subject of future investigation* Incorporate the detailed operational plan in the regional disasterprevention plan for the next year.

(12/12)(8/12)(2/12)

Department in charge
Disaster Countermeasures Division
Disaster Prevention Department
Disaster Countermeasures Division
Disaster Prevention Department
Disaster Prevention Department
Disaster Prevention Department
Disaster Prevention Department
Disaster Prevention Department
Disaster Prevention Department
Disaster Prevention Department

Concluded disaster prevention agreement
no
no
no
no
no
no
no
yes
no
no

(1/12)
Wards inTokyo

Details* Set no plan due to the difficulty in predicting the situation at atime of disaster. Intend to use disaster prevention piers according
to the circumstances when the earthquake occurs.
* Unable to develop the plan because the property of the river may
restrict the navigation of the boats.
* Consider the reception of supplies as a main operation.
* Consider the transportation for the rescue effort.
* Own no boat. Need to request the Tokyo Metropolitan
government to dispatch the boats when the disaster strikes, or to
make an agreement with a shipping company.

3 PUBLIC
AWARENESSOF
DISASTERPREVENTION AND EVACUATION BEHAVIOR(1/12)(5/12)(1/12)(5/12)

2.3 Administration and operation of disaster prevention
piers by river administrators
The administration and operation of disaster prevention piers
under the jurisdiction of the Minisry of Land, Infrastructure,
Transport and Tourism are part of river management. At
present, there is no concrete plan for the administration andoperation of the piers but its formulation is underconsideration. The Arakawa-Karyu Office in the Ministry ofLand, Infrastructure and Transport is considering hosting aworkshop to formulate the rules regarding the operation of thedisaster prevention piers at normal times, but theadministration and the operation of the piers and Inland
navigation facilities remain as an issue. Some disasterprevention piers where municipalities bear the expenses anddraw up the regulations are administered by themunicipalities at normal times and by the national
government in time of an earthquake. The details, however,are not defined.
2.4 Anti-disaster measures based on disaster prevention
plans of disaster prevention piers
The Tokyo metropolitan government has concluded theagreement on transportation between thirteen boat and ship
operators and the Fire and Disaster Management Agency, theBureau of Construction and the Ports and Harbors Bureau, inthe regional disaster prevention plan. However, the detailed
operation remains unclear and is entrusted to the operatorswith full authority. Table 3 indicates that most municipalities
have not concluded a disaster prevention agreement toevacuate the people who are unable to get home by boats. The
government develops partnerships with public transport

3.1 Overview of survey
We conducted an opinion survey of the users of Inland
navigation on their awareness of disaster prevention, as
shown in Table 4.
Table 4: Investigation outline
Item
PeriodType of boat
Fare
Operating hours
MethodNumber of respondents
Navigation route

Summary
August 1, 2008 ~ September 15, 2008
2 electric boats of 10 passengers
Adult: 4,000 yen
4 hours
Interview, Self-completion questionnaire
252 people
Tour around Kanda River, Nihonbashi River and Sumida River

3.2 Need of Inland navigation in time of a disasterFig. 1 shows the result of the survey on the need of Inland
navigation when an earthquake occurs. Seventy-nine percent
of respondents regarded it as necessary and 7 % as
unnecessary, which shows a great need for Inland navigationin time of a disaster. It is considered that people saw thedamaged roads during the Great Hanshin-Awaji Earthquake
and recognized the need of alternative traffic routes.

79

0%

20%
Necessary

40%
Not Necessary

7

60%

80%
Neither

5

9

100%

Other

Figure 1: Recognition of the need of river boattransportation594

3.3 Evacuation behavior in an earthquake disaster
The public awareness of evacuation behavior in an
earthquake disaster is shown in Fig. 2. Thirty-nine percent ofrespondents said that they would walk home; 9 % walk towhere they can take a taxi or a car; 22% take a boat and go
home, if possible; 8% walk to a pier; 3% take a boat and go
home; and 16% go to an evacuation center. The fact thatabout 40% of respondents intended to walk home during anearthquake indicates the low public awareness of theavailability of Inland navigation and disaster prevention piers.
In order to raise the awareness, it is necessary to disseminate
information on the locations of the piers.
39

0%

20%

8

22

8

3

40%
60%
80%
Walk home
Walk to where I can take a taxi or a car
Take a boat and go home, if possible
Walk to a pierTake a boat and go home

16

3

100%

Picture 1: Constructional example

Figure 2: Evacuation behavior in times of earthquake
disasters
3.4 Awareness of using Inland navigation in an
earthquake disasterFig. 3 shows what people expect from Inland navigation
during an earthquake disaster. Thirty-nine percent ofrespondents agreed that it could be used to transport goods;
27% to transport people. Eighteen percent recognized its
function as a hospital ship; 9% as a temporal evacuation
center; and 7% as an accommodation. In order to respond totheir diversified demand shown in the survey and makeeffective use of the boats, a concrete division of roles between
administrative bodies and boat operators is required.

39

0%

20%

18

9

27

40%
60%
80%
Transport goods
Use as a hospital ship
Use as a temporal evacuation centerTransport people
Use as an accommodation

accessible to wheelchair users. A gurney cannot be used totransport sick and injured people and only the transport with astretcher or by foot is possible (see Picture 1).

7

100%

Figure 3: Awareness of using river boat transportation intimes of earthquake disasters
4 POSSIBILITY OF TRANSPORTATION OF SICKAND INJURED PEOPLE
4.1 Problems of the structural design of disasterprevention piers
Previous studies clearly show that the limited use of thedisaster prevention piers at normal times prevents their full
utilization in time of an earthquake disaster. The structures ofthe piers, which building codes are not defined in the
Government Ordinance for Structural Standard for River
Administration Facilities, lack sufficient consideration tovulnerable disaster victims. Many of the piers are not easily

4.2 Disaster base hospitals
The Tokyo metropolitan government has designated the local
disaster base hospitals which are used in time of anearthquake disaster. There are also 152 rear area medical
facilities and one field hospital system which is developed by
the Ground Self-Defense Forces in the property of themunicipalities who own disaster prevention piers. Six piers
are within about 0.5 km as the shortest path from the rear area
medical facilities; 4 are within about 1.0 km. Those 10 piers
are accessible and they account for 14 % of the total.
4.3 Navigability to disaster prevention piers
The boats which can sail to the disaster prevention piers arelimited due to the hull form, properties of the rivers, andfluctuation of physical environment. While all the intended
boats can sail to the piers in Akashicho, the Wakasu Seaside
Park, and Ryogoku, only small passenger boats can navigate
in the Kanda River and the rivers in Koto Ward except thesections, such as the Onagigawa River, because of therestriction (Navigable width of the river > 1.5 × entire lengthof the boat).
According to the Tokyo regional disaster prevention plan,
the procurement of the affiliated boats starts from four days
after the earthquake. If roads are blocked and no traffic isavailable immediately after an earthquake occurs, the waterbuses of the Tokyo Metropolitan Park Association are to be
used as a way to transport medical staffs and severely affected
people. Privately-owned boats such as houseboats are notused in that situation. Since the water buses are not able tonavigate in the Kanda River and the rivers in Koto Warddepending on the tide level and the restriction of the hull form,
small boats are the mainstay of the transportation in an
earthquake.
Floodgates are closed as soon as an earthquake occurs,
which make it impossible to use the waterway transportation
to the disaster prevention piers located in the area below sea
level. It also prevents boats from sailing the area enclosed by
the floodgates.

595

5 CONCLUSION

Figure 4: Navigability
4.4 Possibility of transporting sick and injured people
Table 5 shows the possibility of utilizing the piers intransporting sick and injured people in the initial stage of anearthquake disaster. The Wakasu Seaside Park and the Ariake
Passenger Ship Terminal in the Tokyo Bay area are available
for receiving boats from urban rivers and cooperating withwide-area disaster management bases. The piers in Akashicho,
Ryogoku and Shirahigebashi-south are available as a base of
the banks of the Sumida River. Those five disasterprevention piers are expected to function as a center totransport sick and injured disaster victims.
Other piers have restrictions, such as operations offloodgates, conditions preventing navigation, and danger intransportation routes. Kameido, Tenjinbashi, and Ogibashi
Lock Gate block the navigation due to the closure of the gates.
Although the Yokojikken River and the Kanda River cannot
be used in the initial period of an earthquake because of the
procurement plan of Tokyo, the transportation of many sickand injured people become possible if the small boats which
can sail the narrow rivers are available. The waterway
transportation is accessible in Ichibeigashi if quake resistancein inland areas is enhanced and the fire is contained.
Table 5: Possibility of transporting injured and sickpeople by disaster prevention piers
Pier
Akashi-cho
RyogokuShirahigebashi-south
KameidoTenjinbashi

River

Sumida River

Yokojikken River

Distance (m)Medical facility

270-500
340-360
adjacent
500

St. Luke's International Hospital
Doai Memorial Hospital
Shirahigebashi HospitalTokyo Metropolitan Bokuto Hospital

520

Yujin Hospital

Ichibeigashi

Kanda River

720-1200

Juntendo Hospital, Surugadai Nihon University HospitalOgibashi Lock Gate

Onagi River

960

Asoka Hospital

Ariake Passenger Ship Terminal

Tokyo Bay

700-820

Ariake Hospital

Wakasu Seaside Park

Sunamachi-minami Canal330

A field hopsitalPotential
Obstacle in the initial
Type of boats
Impact on water area danger on thestage of disasterland
All types/

Damaging roadbridges and Damaged
floodgates buildings,
Damaging road Blocked roads
Small boats Securing of boats
bridgesDamaging roadFloodgate operation bridges andfloodgates
Floodgate operation,
Securing of boats

All types///

6 REFERENCES[1][2]

Transportability
Reception from water areas

Damaging roadbridgesThis study shows that there is a need to specify the rules
regarding the administration and operation of disasterprevention piers in regional disaster prevention plans, toenhance public awareness of the piers, and to establish asystem to transport vulnerable disaster victims efficiently.
The concrete plan of administration and operation of theInland navigation should be clearly incorporated in thedisaster prevention plans of the municipalities and that of the
Ministry of Land, Infrastructure, Transport and Tourism.
Based on these observations, we propose the institutional
design as follows.
First of all, a council consisting of administrative bodies
and private-sector entities should be established. It develops asystem for operation, management and cooperation ofdisaster prevention piers. It also consolidates the disasterprevention support system which contains conclusion ofdisaster prevention agreements with private sector entities,
rules of the uniform management of rivers, and a security
system for boat operators.
Second, we propose the development of an operational
plan applicable to disaster base hospitals, disaster prevention
piers, and boats suitable to its water area to construct atransportation network centering on each pier. It enables
smooth and prompt transportation of sick and injured people
in time of an earthquake. It is also important to organize adisaster mitigation system by developing information sharing
system between hospitals and piers, and linking transmission
systems, transportation operations and ship operations.
Additionally, measures to support vulnerable disastervictims in an earthquake, such as development of structural
standards of the disaster prevention piers friendly to theirconditions, should be incorporated in the regional disasterprevention plans. Administrational and operational plans of
the piers should also give attention to this issue. Furthermore,
the system which enables the safe transportation of sick andinjured people should be developed by promoting theaccessibility of the piers at normal times.
The development of these systems is considered to makethe disaster prevention piers safer and enable more efficientuse of the piers in time of an earthquake disaster.

[3]

[4]
Transport to Ariake[5]

Reception from urban rivers

596

Jinnai H, Yoshikawa K and Miura Y, Shuun Toshi: Mizube Kara No
Toshi Saisei. Kajima Shuppankai, 2008.
Egami K, Kondo T, Miura Y and Goto K, “A study on disasterprevention using Inland navigation,” in Proc. of the 63rd Annual
Conference of the Japanese Society of Civil Engineers, 2008.
Koga T, Kondo T, Yamamoto K, Egami K and Orimo R, “A study on
the organization a sea-based relief network system at the time ofepicentral earthquake in metropolitan Tokyo,” The 23rd Conference on
Environmental Information Science, 2009, pp. 101-106.
Egami K, Sakanoi K, Goto K and Ino H, “A Study on Effective Use ofRiver Jetty,” World City Water Forum, 2009, Incheon, Korea, pp.
19-21.
Egami K,Study on Appropriate System Design for Inland Water
Transport and Jetty for Disaster Management. Nihon University
Thesis ,2010

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Model Calculation and Geotechnical Interpretation of the Correlation
Co-efficient Between Finite and Incremental Strain Path of a Formation
J.Rajaraman1 and K.Thiruvenkatasamy 2
but measurement of strain increments is more difficult. Yet,
incremental strain may be more crucial for unraveling thedeformation history of a rock or region than finite strain .The
following figure 1 explains to explore the deformation historyof a rock or region.(from Reference[1]as shown in figure1
the finite strains for the distortion of a square in figure 1a and1b are the same, because the initial and final configuration are
identical.The steps or strain increments by which these final
shapes were reached , however, are very different. We say thatthe strain path of two examples is different , but the finite
strains are the same. The path presented in figure 1a has
incremental strains that reflect a strain ellipse that becomesincreasingly elliptical . In other words , the ratio of the longover the short axis (X/Y)becomes greater . The path in figure
1b, on the other hand, shows that the orientation of the Xi and
Yi axes was perpendicular to those of the finite strain ellipse
during part of the history. Consider this in a geologic context;
the two paths would represent very different strain histories ofa region, yet their finite strains would be identical. Therefore
obviously, an important piece of information is lost without theknowledge of the strain path. It is therefore critical for
structural analysis to distinguish finite strain from incremental
strain.
1.1 Coaxial and Non- Coaxial Strain Accumulation
The principal incremental strain axis rotate relative to the finite
strain axis, is called non-coaxial strain accumulation.The casein which the same material lines remain the principal strain
axis at each increment is called coaxial strain accumulation asshown in figure 2a and 2b .

ABSTRACT: The measure of strain that compares the initial and final
configuration is called the finite strain, which is independent of thedetails of the steps toward the final configuration. When theseintermediate strain steps are determined they are called incremental
strains. The incremental strain may be more crucial for unraveling thedeformation history of a formation than finite strain. In this paper two
different strain sequence are considered for correlation andinterpretation. For both the strain sequences the finite strains reached
are the same. In other words the initial and final configuration are
identical. The steps or strain increments by which these final shapes
were reached, are very different. It means the strain path of two
examples is different, but the finite strains are the same. In one
example the incremental strains reflect a strain ellipse that becomesincreasingly elliptical [the ratio of the long over the short axis [ X/ Y]
becomes greater. In the second example the path shows that theorientation of the X and Y axis was perpendicular to those of the finite
strain ellipse during part of the history. Between these two examples
the correlation co-efficient [r] is calculated. Geotechnically the twopaths would represent very different strain histories of the formation,
yet their finite strains would be identical. If the strain path is notfollowed accurately, many such important piece of information is lost.
Geotechnically it means that the over consolidation ratio [OCR] foreach strain path is different, because the geo fatigue experienced byeach strain path is different. This interpretation is amply supported bycorrelation co-efficient[ r ].

1.0 Strain Path
The measure of strain that compares the initial and final
configuration is called the finite strain ,identified by subscriptf,which is independent of the details of the steps towards the
final configuration. When these intermediate strain steps are
determined they are called incremental strains identified bysubscript i. The summation of all incremental strains (i.e.,theirproduct) therefore , is the finite strain. There are many ways tomeasure finite strain in a rock,

Figure 1The finite strains,Xf and Yf in (a) and (b) are the same,but the strain path by which each wasreached is different,This illustrates The importance of understanding the incremental strain history(Xiand Yi) of rocks and regions and inherent limitations of finite strain analysis.
1,2 Professor, Department of Harbour Engineering and Off Shore
Technology , AMET University, 135 East Coast road, Kanathur -603 112 ,CHENNAI,
INDIA.
E-Mailswamy2667@yahoo.co.in1
rajaraman.usha44@gmail.com2

597

Figure 2 Non-coaxial [a] and coaxial [b] strain. The
2.2 CORRELATION COEFFICIENT ( r )incremental strain axes are different material lines during
Let (x1 ,y1) , (x2, y2) … (xn ,yn) be a given set of n pairs of
non-coaxial strain. In coaxial strain the incremental strain axes observation on two variables x and y. The correlation coefficientare parallel to the same material lines. Note that the magnitude or coefficient of correlation , between x and y (denoted by thesymbol r ) is then defined asof the strain axes changes with each step. From [1].
,r=where σx and σy are the standard derivation of x
2. CORRELATIONandThe word “correlation” is used to denote the degree ofy respectively cov ( x,y) denotes the covariance of x and y
association between variables. If two variables x and y are so(section
9.6). This expression is known as Pearson’s product –related that variations in the magnitude of one variable tend to
moment formula , and is used as a measure of linear correlation
be accompanied by variations in the magnitude of the othervariable, they are said to be correlated. If y tends to increase as x between x and y.The formula for r may be written in various
increases , the variables are said to be positively correlated. If y other forms. Putting the explicit expression for cov(x,y) , σx and σydecreases as x increases , the variables are negatively correlated in (9.7.1), and multiplying both the numerator and the
If the values of y are not affected by changes in the values of x , denominator by n , we have∑the variables are said to be uncorrelated. Correlation may alsor=∑.
be linear or non-linear. If the amount of change in one variable
tends to bear a constant ratio to the amount of change in the
now expanding the expressions
other variable , then correlation is said to be “linear”;because the∑scatter diagram would show a linear path. Here we shall ber=concerned with linear correlation or simple correlation only.
∑ ²
²
² . ∑
This is measured by “Correlation Coefficient”
2.1 COVARIANCE
Given a set of n pairs of observations (x1 , y1 ), (x2,y2),…..(xn ,yn)relating to two variables x and y , the covariance of x and y ,usually represented by cov ( x, y) , is defined as(y- )cov (x,y) = ∑expanding the equation on the right, it can be shown that∑∑∑cov ( x, y) =this is generally used for calculation as in [2].

Multiplying the numerator and the denominator by n again
and since nx = ∑x and n =∑y, we may write r
∑
∑
∑
r=
∑

∑

² .

∑

∑

²

Note : In all the forms shown above , the denominators
contains two factors under the square root. They may be obtain
On replacing y by x , and x by y in the numerator. as in [2].

598TABLE I Correlation Coefficient from strain ratios (x/y)
from figure 1.

2.3 Properties of Correlation Coefficient:(i) The correlation coefficient r is independent of the choice ofboth origin and scale of observations. This means that ifćand v =
U=ď
Where c,ć ,d ,ď are arbitrary constants (d and ď positive ), thenr xy = ruv i.e. correlation coefficient between x and y= correlation coefficient between u and v
In general , if X = a + bx , Y = á + b` y, thenr xy = ± rxyaccording as b and b` have the same sign or opposite signs.
(ii) The correlation coefficient r is a pure number and isindependent of the units of measurement. This means that if, forexample x represents height in inches and y weight in lbs., thenthe correlation coefficient between x and y will neither be in
inches nor in lbs. or any other unit , but only in number.
(iii) The correlation coefficient r lies between -1 and +1 ; i.e. r
cannot exceed 1 numerically.
-1 r +1
²

=
²

∑ ²

--

∑

.

.

=
∑

.

= (7.656) – (3.1) (2.36)= 7.656 – 7.316= 0.34

.

,

r=
=

.

ₓ
.
.

.

.
.

√ .

.

= 0.6083

2.4 Strain History And Correlation Coefficient ( r )(flexibility of r as a method ).
The strain ratios need not necessarily be small values .The
strain ratio can be larger in size. The statistical methods canhandle different types of strain ratios. This is illustrated by astatistical method which is also applicable to strain ratios( coaxial stains ) The general data assumed to find thecoefficient of correlation is given below :².

TABLE II Illustrative example for x,y.
X
65
63
67
64
68

62

70

66

Y

66

68

65

68

66

68

65

69

Solution : Since the correlation coefficient is unaffected bychange of origin ( and also scale ), let us change the origins of
X and Y to 65 and 67 respectively so we can write x = X – 65 ,y = Y – 67 .The new table for calculations for correlation
coefficient is as follows :

²

∑

.

²
.

²

²

.

.

=

.

²

.

=

.

Covariance (x, y) =²

= 6.2 – (2.36)2= 6.2 – 5.569 = 0.631

599

TABLE III Correlation Coefficient ( r) data
X
( 1)
65
63
67
64
68
62
70
66
525

Y
( 2)
68
66
68
65
69
66
68
65
535

x=X-65
( 3)
0
-2
+2
-1
+3
-3
+5
+1
5

y=Y-67
( 4)
1
-1
1
-2
2
-1
1
-2
-1

x2
( 5)
0
4
4
1
9
9
25
1
53

Y2
( 6)
1
1
1
4
4
1
1
4
17

xy
7
0
2
2
2
6
3
5
-2
18

²
ₓ =

-

²
ʏ =

-

²=
²=

Cov (x.y) =
,
ₓ ʏ

.65

Therefore, the required correlation coefficient is .64
The origin is changed to values near the means. The means of
X and Y are 525/8 = 65.625 and 535/8 = 66.875 respectively.

Conclusions :1. In a geologic context two different strain paths maybe
very different in strain histories of a region yet theirfinite strains would be identical .without this if thefinite strain analysis is carried out the correlation
coefficient are will be unity or one.
2. In reality the correlation coefficient is only 0.6083
3. If an important strain history is lost and yet we getcorrelation coefficient as 1 is erroneous .
4. The correlation coefficient r is flexible and theproperties of r is capable of managing large data byshifting the origin as stated in illustration 2.
5. The deviation of correlation coefficient from 1
indicates different Geo – fatigue due to incremental
strain during stress history of the region.
6. By preparing correlation coefficient data for many
Geo-profiles the stress history through strains could
be prepared.
7. Rapid change in strain data / correlation coefficient
will indicate the complex Geology especially inJapan.

References
1.
2.
3.

600

Van der pluijm, ben a, Earth Structure-2003
W.W.Norton & company inc, New York 10110.
pp65-68
Das N.G.Statistical Methods combined edition
vol.1&2 2009 Tata McGraw-Hill co New Delhi
pp275 -280
Jensen Jerry,1997,Statistics for petroleum engineersand geoscientists, Prentice Hall, Cuevas 1993 p32 -36

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Geocell Reinforced Bed Subjected to Simulated Traffic Loads
P. Sharifi, S.N. Moghaddas Tafreshi., Department of Civil Engineering, K.N. Toosi University of Technology,
Tehran, Iran. E-mail: Sharifipayam.civil@gmail.com



ABSTRACT: This paper describes a series of laboratory, pilot scale
tests were performed on geocell reinforced sand beds with different
number of geocell layers and with different heights of geocell interms of H/D (H: the thickness of geocell and D: the loading surface
diameter) under repeated loads to simulate the vehicle traffic loads.
The influences of the thickness of geocell layers and arrangementsof geocell layers embedded in bed were studied. Assessment of
performance was investigated for different number of layers while
the used mass of geocell material, was kept constant. For example,
the experiment reinforced by two layers of geocell with H/D=0.5 hasexactly the same mass of geocell compared with the experiment
reinforced by one thicker layer of geocell reinforcement with
H/D=1.0. The results show that the reinforced bed with more layersof geocell has more efficiency than one layer of geocell in reducing
the surface settlement under repeated loading while the mass of
geocell used in the foundation bed is kept constant.

different factors on the soil-geocell behavior such as numberof geocell layers.
2. TESTING EQUIPMENTS
The testing apparatus is comprised of three main parts include
the loading system, the testing tank and the data acquisition
system. The general arrangement of the laboratory test is
shown in Fig.1.

Keywords: geocell-reinforced sand; soil surface settlement,
repeated load, geocell layers
1. INTRODUCTION
In recent decades, due to its economy, ease of construction
and ability to improve the visual appearance, reinforced soil
has been widely exploited in geotechnical engineering
applications such as the construction of roads, railway,
stabilization of slopes, improvement of soft ground,embankments and footings {(Das and Shin, 1994); (Raymond,2002); (Shin et al., 2002); (Hufenus et al., 2006); (Dash et al.,2004); (Sitharam et al., 2007); (Al-Qadi et al., 2007);(Moghaddas Tafreshi and Khalaj, 2008); (Sireesh et al.,2009); (Moghaddas Tafreshi and Dawson, 2010a);(Moghaddas Tafreshi and Dawson, 2010b)}. Generally, whena reinforced soil is subjected to loading, it shows a higher load
carrying capacity and a lower settlement compared with a soil
mass without reinforcement {e.g., (Dash et al., 2003); (Yoonet al., 2004); (Patra et al., 2005)}. Among of the different
types of geosynthetics-reinforcement, in the case of
geocell-reinforcement, in addition to the tensile strength of
the reinforcement and friction at the soil-geosynthetic
interface, a confining stress will be developed on the infillingsoil due to passive resistance provided by the cell structure{(Bathurst and Karpurapu, 1993); (Madhavi et al., 2006)}
Thus, the lateral resistance in the geosynthetic restrains thelateral movement of the soil and, consequently, reduces thevertical deformation. In this paper, an experimental study
wasconducted
tounderstand
the
role
of
geocell-reinforcement to mitigate the hazardous effect of
repeated loading. A series of tests on the single and double
layers of geocell were performed to evaluate the role ofFig. 1. General arrangements of tests.

The loading system is composed of a loading frame, ahydraulic actuator and a controlling unit. The loading frame
includes two stiff and heavy steel columns of 1600 mm height
and a horizontal beam of 1270 mm length, which supports thehydraulic cylinder. The controlling unit is comprised of anelectro-mechanical valve, which can regulate the intensity of
the compressed oil required to produce a cyclic load with thedesired amplitude and frequency. The soil beds were preparedin a test tank (rigid steel box) with inside dimensions of
700mm in length, 700mm in width, and 600mm in height.
According to the preliminary tests, the measured deflection of
the side faces of the tank proved to be negligible and in theranges to satisfy the rigidity of the system. The geometry of
the test configurations for the geocell-reinforced bed is shownin Fig.2. In order to provide experimental control andrepeatability of the tests, the compaction energy produces bymeans of hydraulic cylinder which applies constant pressure
on a wooden stiff plate (680 mm * 680 mm). The wooden
plate is fitted to the soil surface so all the energy will transfer
to soil uniformly. A special data acquisition system wasdeveloped by which stress and settlement could be read and601

recorded automatically. The system was able to read the datafrom channels simultaneously. An S-shape load cell was also
used and placed in the loading shaft to measure the pattern of
the applied loads on the trench surface accurately. Two
LVDTs were placed on the loading surface to measure thesettlement of soil surface during the repeated loads. The
loading surface was simulated by a rigid steel plate andmeasured 110mm diameter (D) and 30mm thickness.

Table 1. The engineering properties of the Geotextile used in thetests.
Description
Value
Type of geotextile
Non-woven polymer
Type of polymer
100% polypropylene
Area weight (g/m2)
190
Thickness under 2 kN/m2 (mm)0.57
Thickness under 200 kN/m2 (mm)0.47
Tensile strength (kN/m)
13.1
Strength at 5% (kN/m)
5.7

4. PREPARATION OF MODEL TEST

Fig. 2. Geometry of geocell-reinforced foundation bed.

3.

TESTING MATERIALS

According to Fig. 2, the soil used in the tests was natural soil
with grains size between 0.002 and 15 mm, D50=1.4 mm,
Cc=1.2, Cu=13.86 and Gs=2.645. This soil is classified as SW
in unified soil classification system. The maximum andminimum porosities of this soil (emax and emin) were obtained
0.66 and 0.36, respectively. The relative density of soil was
selected 70% in all of the tests. Geocells used asreinforcement was made of a type of non-woven Geotextile –an innovative approach for use in ground stabilization. Whencells fill with soil, the combination of geocell and soil
provides an ideal surface for construction projects such aspavement, road, foundations, slopes, driveways and so on.
The pocket size of the geocell kept constant of 50 mm). It wasused at heights (H) of 25, 50, and 100 mm in the testing
program. Fig. 3 shows an isometric view of the geocell usedin the investigations. The engineering properties of thisgeotextile, as listed by the manufacture, are presented inTable 1.

Fig. 3. Isometric view of the geocell

At first the piston force and number of impact to achieve thedesired density was determined by a series of trials withdifferent piston force and number of impact. Sand was thencompacted in layers of specific thickness depend on height of
the geocell used to consistently maintain a relative density of
70% in all the tests. According to the test type, each of thegeocell-reinforced layer was placed carefully and prior tostart compaction a surcharge approximately 10mm wasapplied above geocell-reinforced matters. Soil mixture placed
either between the two geocell layers or inside the pocket of
geocell. The surface of the soil was then leveled and the steel
plate as a model of loading surface was centered in the testingtank. A load cell was placed in the loading shaft to record theapplied loads and two LVDTs were placed on the steel plate
accurately to measure the settlement of soil surface during therepeated loads. Reference {(Moghaddas Tafreshi andDawson, 2010b)} reported that the optimum distance
between of the topmost layer of the geocell and soil surface
under repeated load should be approximately 0.1 times of theloading surface width. Hence, in this investigate in all tests
the first layer of reinforcement was placed at u/D=0.1.
5.

PATTERN OF APPLIED LOAD

Trapezoidal load was used by many researchers for modelingtraffic load and the loading arrangements were those
representatives of typical trucks on LVRs in the United
Kingdom. Although lorries may run with tyre pressures as
low as 240kPa and as high as 900kPa, pressures of 400kPa
and 800kPa were deemed to be closer to the operational
values {(Brito et al., 2009)}.

Fig. 4 Typical time history of initial static and repeated load on
footing.

602

Fig. 4 shows the typical time history of applied load on thesoil surface used in present experiments. As can be seen, thesoil surface is subjected to a trapezoidal load having theminimum amplitude of 0 kPa and the maximum amplitude of
400 kPa. A trapezoidal load cycles with a frequency of 0.5 Hz
would be continued until the rate of change of total settlement
drops to an insignificant amount or, alternatively, excessive
settlement and unstable behavior is observed. In the case of
stable tests, settlement due to up to 20,000 subsequent load
repetitions was recorded.

6.

TEST PARAMETERS AND TESTING PROGRAM

14 tests in different series were planned and carried out in thisresearch. According to table 2, the tests were scheduled tofind out the effect of the distance between of the first geocell
layer and second geocell layer (h), the number of geocell
layers (N) and performance of a composite material(geocell-reinforced soil) on the behavior of foundation bed
under cyclic load. All these variable parameters used todescribe the tests are expressed in non-dimensional form withrespect to loading surface width (D) as H/D and h/D. The
value of b/D was kept constant 3.2 in all reinforced tests{(Moghaddas Tafreshi and Dawson, 2010a); (Moghaddas
Tafreshi and Dawson, 2010b)}. It should be noted that someof the tests were repeated carefully at least twice to examine
the performance of the apparatus, the accuracy of themeasurements, the repeatability of the system, reliability of
the results and finally to verify the consistency of the test data.
The results obtained depicted a close match between resultsof the two or three trial tests with maximum differences inresults of around 10%. This difference was considered to besmall and is subsequently neglected. It demonstrates that theprocedure and technique adopted can produce repeatable tests
within the bounds that may be expected from geotechnicaltesting apparatuses.
Table2. Tests schedule

Type ofreinforcement

unreinforcement
geocell
reinforcement

7.

In order to show the influence of the thickness of sandbetween two layers (i.e., the distance between two subsequentgeocell layers: ratio h/D), two test series was done. Fig. 5
summarizes the variation of the footing settlement(non-dimensional parameter as S/D) with time for the two
geocell-reinforced beds (N=2) with a constant thickness of
geocell (H/D= 0.45) and two different h/B values of 0.1 and
0.2 (h/D = 0.1, 0.2) and also for unreinforced soil bed. The
results illustrated in Fig. 5 show that in the case of geoecell
reinforced beds, the rate of change of both the total settlement
and accumulated plastic settlement reduces as the number ofcycles (or time) increases.

Numberofgeocelllayers, N
-1, 2
2
1

H/D

h/D

No. of
Tests

-0.45
0.25
0.9

--0.1, 0.2
--

1+1*
2+2*
2+2*
1+1*

Fig. 5. Variation of settlement with time for unreinforced andreinforced beds (N=2, H/D=0.45 and h/D=0.1 and 0.2).

The variation of settlement becomes stable after about 18000
number of load cycles. Besides, Fig. 5 indicates that in thecase of unreinforced bed, after the initial load cycles (about
10 in this case), settlement of the footing increases
significantly and then excessive settlement and, consequently,
unstable behaviour is observed.
The results depict for the reinforced beds the total footing
settlement is considerably decreased relative to theunreinforced settlement. It is also clear that the rate of
reduction in the total footing settlement reduces with increasein the value of h/D as no marked further decrease in footing
settlement may be expected when the distance between two
geocell layers, h/D increases to h/D>0.2.
7.2.
COMPARISON OF SINGLE AND TWINLAYERS OF GEOCELL REINFORCEMENT AT THE
SAME MASS OF GEOTEXTILE MATERIAL USEDRESULTS AND DISCUSSION

In this section, the tests results of the laboratory model arepresented with a discussion highlighting the effects ofinvestigated parameters.
7.1. THE EFFECT OF THE THICKNESS OF SAND
BETWEEN TWO SUBSEQUENT GEOCELL LAYERS(h/D)

The performance of single layer and twin layers of
geocell-reinforcement, for the same mass of geotextile
material used, in decreasing the settlement of a sand bedis the
subject of Fig. 6 and Fig. 7. The variation of the footing
settlement (in terms of S/D) with time (or number of load
cycles) as a function of the number of layers of geocell (N)and thickness of geocell (H/D) as compared withunreinforced soil bed is shown in Fig. 6 and Fig. 7.

603

Overall, these figures indicate that for the same mass ofgeotextile material used, the value of maximum footing
settlement decreases considerably due to more layers of
geocell (N). On the other hand, the geocell-reinforced bed
system with 2 layers clearly has a superior ability to reducethe maximum settlement of the footing as compared with thegeocell-reinforced bed system with 1 layer system.
For example, the maximum footing settlements for the 2layers of geocell-reinforced soil bed with H/D=0.225 (h/D=0.2) and for one layer of geocell-reinforced soil bed with
H/D=0.45 at the end of loading are36% and 28% of theloading plate diameter, respectively (as per Fig. 6).

consequently using more geocell layers with lower thickness(while the mass of geocell is kept constant) would be more
beneficial, (2) if the reinforced zone is considered analogous
to a beam, the competent thickness of soil layer between twolayers increases the moment of inertia, and hence shear andbending rigidity of the geocell zone significantly increases
Therefore, it would be expected the multiple geocell-sandmattress has more effective than one layer with the samemass.
The hysteresis loop of loading plate settlement for theunreinforced bed, the 2 layers of geocell-reinforced soil bed
with H/D=0.225 (h/D= 0.2) and one layer of
geocell-reinforced soil bed with H/D=0.45, shown in Fig.8.
The hysteresis curve shows that there is an increase in theslope of the repeated pressure-settlement curve with anincrease in the number of load–unload cycles. This
corresponds to a reduction in the rate of change of peaksettlement with number of load applications which, together,
are indicative of a hardening of the reinforced system. The
general behavioural patterns observed in these tests is in-linewith those observed in the repeated load testing of
unreinforced and reinforced bed as observed by several
authors {(Moghaddas Tafreshi and Dawson, 2010b);(Werkmeister, 2005)}.

Fig. 6. Variation of settlement with time for unreinforced andreinforced beds (N=2,H/D=0.225 with h/D=0.1 and N=1,
H/D=0.45)

Fig. 8. Hysteresis curve for unreinforced and reinforced beds
(N=2, H/D=0.45 with h/D=0.2 and N=1, H/D=0.9).

8. SUMMERY AND CONCLUSIONS

Fig. 7, Variation of settlement with time for unreinforced andreinforced beds. (N=2, H/D=0.45 with h/D=0.2 and N=1,
H/D=0.9)

The increase in performance improvement of two layers ascompared with one layer could be attributed to two following
reasons (1) the thicker geocell may be made a major obstacle
to achieve a suitable density inside of geocell's cells,

A series of laboratory tests have been carried out on theunreinforced and geocell-reinforced bed which is a
representative of traffic load on the surface road. Benefits
were assessed in terms of the decreased settlement of a soil
surface subjected to the repeated loads. The various
parameters studied in this testing program include: number ofthe geocell-sand mattress and thickness of sand between
layers. The repeated load test results show that the rate ofchange in peak settlement and residual settlement reducedwith the cycle numbers increasing. Also as the number of
geocell layers increase, the settlement of the foundation bed
decreased. Furthermore, the performance of the604

geo-composite material increases when the number ofreinforced layer increases, while the mass of geocell is kept
constant. For example, the settlement for N=1 (H/D=0.45) isapproximately 28% more than the settlement with N=2(H/D=0.225 and h/D=0.2) at the same mass of geocell.

[17] Werkmeister, S, Dawson, A R & Wellner, F., 2005. Permanent
deformation behavior of granular materials: the shakedown
theory. Transportation Research Board. 6 (1), 31-57.
[18] Yoon, Y.W., Cheon, S.H., Kang, D.S., 2004. Bearing capacity
and settlement of tire-reinforced sands. Geotextile andGeomembranes 22 (4), 439–453.

9.

[19] beds. Geotech Geol Eng, 25:509–524.

REFERENCES[1] Al-Qadi, I.L., et al., 2007. Accelerated full-scale testing ofgeogrid-reinforced flexible pavements. TRB 2007 Annual
Meeting (CD-ROM). Washington, DC: Transportation Research
Board, National Research Council.
[2] Bathurst, R.J., Karpurapu, R., 1993. Large-scale triaxial
compression testing of geocell- reinforced granular soils.
Geotechnical Testing Journal, ASTM 16, 296–303.
[3] Brito, L.A.T., Dawson, A.R., Kolisoja, P.J. 2009. Analytical
Evaluation of Unbound Granular Layers in Regard to Permanent
Deformation. In: Proceedings of the 8th International on the
Bearing Capacity of Roads, Railways, and Airfields (BCR2A’09),
Champaign IL, USA, pp 187-196.
[4] Das, B.M., Shin, E.C., 1994. Strip foundation on
geogrid-reinforced clay: behavior under cyclic loading.
Geotextiles and Geomembranes 13 (10), 657–667.
[5] Dash, S.K., Rajagopal, K., Krishnaswamy, N.R., 2004.
Performance of different geosynthetic reinforcement materials insand foundations. Geosynthetics International 11 (1), 35–42.
[6] Dash, S. K., Sireesh, S. and Sitharam, T. G. (2003). Behaviour of
geocell reinforced sand beds under circular footing. Ground
Improvement, 7, No. 3,111–115.
[7] Hufenus, R., Rueegger, R., Banjac, R., Mayor, P., Springman, S.M.,
Bronnimann, R., 2006. Full-scale field tests on geosynthetic
reinforced unpaved on soft subgrade. Geotextiles andGeomembranes 24 (6), 21–37.
[8] Madhavi Latha G., Rajagopal K. and Krishnaswamy N.R. 2006.
Experimental and theoretical investigations on geocell supported
embankments. Int. J. of Geomechanics, ASCE, 6(1),[9] Moghaddas Tafreshi, S. N. and Dawson, A.R. 2010a. Behavior offootings on reinforced sand subjected to repeated loading –Comparing use of 3D and planar geotextile. Geotextiles andGeomembranes, Elsevier, 28, 434–447.
[10] Moghaddas Tafreshi, S. N. and Dawson, A.R. 2010b.
Comparison of bearing capacity of a strip footing on sand with
geocell and with planar forms of geotextile reinforcement.
Geotextiles and Geomembranes, Elsevier, 28, 72–84.
[11] Moghaddas Tafreshi, S.N., Khalaj, O., 2008. Laboratory tests ofsmall-diameter HDPE pipes buried in reinforced sand under
repeated load. Geotextiles and Geomembranes 26 (2), 145–163.
[12] Patra, C.R., Das, B.M., Atalar, C., 2005. Bearing capacity ofembedded strip foundation on geogrid reinforced sand. Geotextile
and Geomembranes 23 (5), 454–462.
[13] Raymond, G.P., 2002. Reinforced ballast behaviour subjected torepeated load Geotextiles and Geomembranes 20 (1), 39–61.
[14] Shin, E.C., Kim, D.H., Das, B.M., 2002. Geogrid-reinforced
railroad bed settlement due to cyclic load. Geotechnical andGeological Engineering 20 (3), 261–271.
[15] Sireesh, S., Sitharam, T.G., Dash, S.K, 2009. Bearing capacity ofcircular footing on geocell–sand mattress overlying clay bed with
void. Geotextiles and Geomembranes, Elsevier.
[16] Sitharam G., Sireesh, S. T. and Dash, S. K. 2007. Performance of
surface footing on geocell-reinforced soft clay beds. Geotech
Geol Eng, 25:509–524.

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First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21‐23, 2011, ISBN: 978‐4‐9905958‐0‐7 C3051

Effects of Gradation of Various Sand Deposits of Pakistan on
Strength of Hardened Concrete
Gauhar Sabih 1 and S. Muhammad Jamil2
1

Graduate Student, School of Civil and Environmental Engineering, National University of Scienceand Technology, Islamabad, Pakistan, gauhar73@yahoo.com
2
HOD, Department of Geotechnical Engineering, National University of Science and Technology,
Islamabad, Pakistan, smuhammadjamil@ymail.com

ABSTRACT: Various sand deposits of varying gradation areavailable in different regions of Pakistan. The gradation ofdifferent sands was compared with the ASTM specifications. It
was revealed that only two sands satisfy the ASTM gradation
limits. The same was also confirmed by the results ofworkability and compressive strength. For betterment of theconstruction industry, optimization of sand was carried out. The
grading curves of optimized sands became compatible with the
ASTM specifications and the concrete made with improved
sands showed up to 39% increase in compressive strength. The
optimized sands were cost effective also.

various types of sands in the light of ASTM standards andto contrast the effects of gradation on the strength andworkability of concrete. Finally devising a technique for
optimizing the gradation of fine sands by sieving thestone dust and mixing different fractions of stone dustwith the natural sand to get optimized sand gradation andevaluating its effects on strength of hardened concrete.
2. IDENTIFICATION OF SAND DEPOSITS
Various well known material suppliers and construction
firms were approached for identification of major sanddeposits of the country. The locations of identified sanddeposits are shown in Fig 1.

Key words: sand, concrete, Pakistan, Lawrencepur, stone dust,
compressive strength

1. INTRODUCTION

3. EXPERIMENTAL PROGRAM

Since up to 30 percent of the total volume of concreteconsists of sand/fine aggregate consequently fineaggregate characteristics significantly affect theperformance of fresh and hardened concrete and have an
impact on the cost effectiveness of concrete [1].
Gradation or particle size distribution of sand affects theproperties of concrete like packing density, voids content,
workability and strength [2]. Uniformly distributed
mixtures generally lead to higher packing resulting in
concrete with higher density and less permeability, andimproved abrasion resistance. Consequently, uniformly
distributed mixtures require less paste, thus decreasing
bleeding, creep, and shrinkage [3]-[4]. The fine aggregate
should be uniformly graded. If fine aggregate is toocoarse it will produce bleeding, segregation and harshness,but if it is too fine, the demand for water will be increased[5].
Various sand deposits of varying gradation are available
in different regions of Pakistan. The existence of large
water reservoirs usually transports finer sands towards
downstream; consequently the sand in the Northern
region is coarser as compared to the sand from Central
and Southern regions. There is a general trend in the
country to use coarser sand from Lawrencepur, a site
located in the Northern region. Transportation cost ofsand from this place to far flung areas in Centre and Southmakes any project cost intensive. This study was aimed atevaluating the grain size distribution and absorption of

3.1 Sieve Analysis and Classification

607

All the sand samples were classified by sieve analysis,
carried out according to ASTM C-136-96 [6] and theclassification of sand was done according to ASTM D2487-93 [7]. ASTM C-33-03 specifies the gradation limits
for sands to be used as fine aggregate in concreting [8].
The gradation curves of all the sands were compared with
the ASTM gradation limits to contrast the variations.
3.2 Absorption Percentage
All the sand samples were tested for their absorption
percentage in saturated surface dry (SSD) condition
according to ASTM C-128.
3.3 Organic Impurities in Sands
This test was conducted for all the sands according to
ASTM C-40 [9]. The test procedure consisted of making areference standard colour solution by dissolving reagent
grade potassium dichromate in concentrated sulfuric acidat the rate of 0.25 gram per 100 ml of acid. The sandsample was filled in a glass bottle up to 130 ml mark andsodium hydro oxide solution (3% by mass in 97% water)
was added up to 200 ml mark. The bottle was shaken
vigorously and allowed to stand for 24 hours. After 24hours the colour of the sand solution was compared withfreshly made reference colour solution. The darker colour

Fig 1. Locations of Identified Sand Deposits

than reference colour indicates the possibility of sandcontaining injurious organic impurities and requires
further testing for sand before use in concreting. The
lighter or equal colour means that the sand does notcontain injurious organic impurities and no further testingis required.
3.4

4.

4.1 Sieve Analysis and Classification
All the sand types are poorly graded sands classified as
SP, except the Wadd sand which contains silt as well andclassified as SP-SM. The Percentage of micro fines(material passing No. 200 sieve) is below 5% in all sands
except Wadd sand which contains 9% micro fines.
According to ASTM C-33, the allowable limit of microfines in sand for concreting is up to 5% so all the sands
satisfy the ASTM criteria except the Wadd sand.

Evaluation of Sands in Concreting

3.4.1 Concrete Mix
The concrete mixes were prepared using sand / fineaggregate from eight different sources and remaining
variables as constant with the cement: sand: coarse
aggregate ratio of 1: 2: 4 (by weight) respectively, cement
content (201.7 kg/cum) and water cement ratio of 0.5 withconstant slump value. The course and fine aggregateswere used in SSD condition. The course aggregate used
was taken from Margallah quarry located in the Northern
region. The cement used was Fauji cement (type 1,
portland cement). The water used was normal tap water
without any undesired colour.
3.4.2

TEST RESULTS AND DISCUSSIONS

4.2 Comparison of Gradation with ASTM Gradation
Limits
The comparative results of all the sands with the ASTM
gradation limits are shown graphically in Fig 2. All the
sands are finer than ASTM gradation limits except the
Qiblabandi and Bollari sand. The Qiblabandi sand is thebest match to ASTM gradation limits while Bollari sandis little finer than Qiblabandi sand.

Workability of Fresh Concrete

The workability was evaluated by measuring the slump offresh concrete according to ASTM C-143. The concrete
was filled in the slump cone in 3 equal layers with 25
blows of tamping rod each. After leveling the top layer,
the cone was lifted and slump was measured as reductionin original height.
3.4.3 Compressive Strength of Hardened Concrete
The test was conducted on 4x4 inches cubes (as per BS1881) and 4x8 inches cylinders (as per ASTM C-39). The
compressive strength was determined at the age of 3, 7,
14 and 28 days [10]-[11].

608thus determining the improvement in the compressive
strength of concrete specimens.

Fig 2. Sand Gradation versus ASTM Gradation Limits
4.3 Absorption Percentage
All the sands have absorption less than 3 % so they arewithin range of the limiting range as per BS-8007. The
Qiblabandi sand has the lowest absorption percentage i.e.
1.75% being the coarsest sand of all types while the
Bollari sand has the highest absorption percentage of
2.11% but it is also within range of the acceptable limit.
4.4 Organic Impurities in Sands
All the sand sample solutions have the colour lighter than
the reference standard colour solution so all the sands do
not contain injurious organic impurities and no further
testing is required for the organic impurities (according to
ASTM C-40). Consequently all the sands are suitable forconcreting as per the results of this test.

Fig 3. Comparison of Compressive Strength ofConcrete Cylinders (psi)

4.5 Workability of Fresh ConcreteThis resulted in significant improvement of sandgradation as well as the compressive strength of concrete.
Six sands which were out of range of ASTM standards
were optimized. Various different combinations weretried for making the improved sand samples to attain thebest mix.

The slump values of the concrete made with different
types of sands are ranging between 1.1 to 1.2 inches withsame water cement ratio. This slump value was taken as
constant for further compressive strength testing of
hardened concrete. According to specifications, handtamping was resorted in the preparation of cubes while
vibrator insertions were used in preparation of cylinders[12]-[13].

5.1 Gradation of Optimized Sands
The gradation curves of optimized sands were comparedwith the gradation curves of original sands in the light of
ASTM gradation limits to determine the effects on thegradation curves. The comparison clearly shows that thegradation curves of optimized sands are satisfying the
ASTM gradation limits where as the original ones wereout of the limits. The comparison of original andoptimized gradation curves is shown in Fig 4.

4.6 Compressive Strength of Concrete Cylinders andCubes
The compressive strengths of both of concrete cubes andcylinders follow the same trend line thus confirming theresults. The comparative results of compressive strengthof concrete cylinders are graphically shown in Fig 3. The
Qiblabandi sand proved to be the best fine aggregate
giving highest compressive strength of 3463 psi.
Gradation meeting the ASTM specifications, lowest
percentage of micro fines and within range finenessmodulus of Qiblabandi sand are the supporting factors forthe highest compressive strength. The trend of gradationof different sands as compared with the ASTM gradation
limits is also confirmed by the results of the compressive
strength. The Ghazi, Ravi and Chanab sands are below
3000 psi which is not desirable for concreting works.

5.2 Compressive Strength of Concrete
The optimized sand samples were tested by making
concrete cubes and cylinders with the cement: sand:coarse aggregate mix of 1: 2: 4 (by weight) and thevariables like cement (201 kg/cum), water-cement ratio(0.5), curing conditions and course aggregate (Margallah)
as constant. The concrete specimens were than tested forcompressive strength at the age of 3, 7, 14 and 28 days to
get the improved compressive strengths. These improved
strengths were than compared with the originalcompressive strength values to get the percentage
improvement.

5. OPTIMIZATION OF SAND GRADATION
Having observed the gradation of various sands, anendeavor was made to bring them within the specified
limits of ASTM by adding a fraction of coarse material.
The stone dust from Margallah quarry site was taken andvarious fractions of stone dust were produced by sieving
the stone dust on Number 4, 8, 16, 30 and 50 sieves(These fractions are denoted as R4, R8, R16, R30 and
R50 in this study). Different percentages of various
fractions of stone dust were added to the sand samples
and optimized samples were produced. The optimized
samples were than tested for their gradation to check thecompatibility with ASTM limits. The optimized samples
were then used for making concrete cubes and cylinders

609

The comparative summary is shown graphically in Fig 6.
The comparison of percentage improvement of optimized
sands in comparison to original sands is shown in Fig 7.
The results show that all the optimized sands gaveincreased compressive strength. The optimized Ravi andChanab sands show the most significant increase of 37%
to 39% in the compressive strength when compared withoriginal sands.
7. CONCLUDING REMARKS
Based on laboratory investigations and data analyses,
following conclusions are drawn:

Fig 4. Comparison of Optimized Gradation Curves
with ASTM Limits

5.3 Results of Compressive Strength of Optimized
Sands



The compressive strengths of both cubes and cylinders
follow the same trend line thus confirming the results.
The comparative results of compressive strength of
concrete cylinders are graphically shown in Fig 5. The
results show that there is significant improvement in thecompressive strengths and all the improved sands areabove 3000 psi which is the foremost requirement for asand to be used in concreting works. The improved
Lawrencepur sand and Durrani sand (with finenessmodulus of 2.62 and 2.58) respectively are giving thehighest strength of 3441 psi and 3547 psi respectively.
6.





COMPARISONOFORIGINALAND
OPTIMIZED SAND PROPERTIES BASEDON COMPRESSIVE STRENGTH OF
CONCRETE

Local sands can be utilized in the construction
projects by optimizing the gradation using stone
dust to avoid high costs of transportation.
Even the best considered sand deposit requires
fractions of stone dust to maximize its potential
in concreting.
The optimized Ravi and Chanab sands show themost significant improvement of 37% to 39% in
the compressive strength while the optimized
Lawrencepur, Ghazi, Wadd and Durrani sands
also show greater strength than original sandswith percentage improvement of 8% to 16%.
The optimized Ghazi, Ravi and Chanab sands
can be utilized in concreting works where as they
were not suitable for concreting in their originalform.
All the optimized sands show higher
compressive strength than the original sands inconcrete cubes and cylinders both and are above
3000 psi which is the minimum criteria for any
sand to be used as fine aggregate in concreting
works.

The cylindrical compressive strengths of optimized sands
at the age of 28 days were than compared with thecylindrical compressive strength values of original sands
to get the percentage improvement.

Fig 6. Comparison of Original versus Improved
Cylinder Strength

Fig 5. Comparison of Improved Compressive strengthof Cylinders (psi)
6.1 Comparative Results of Cylinder Strength610

[11] British Standards Institution, "Methods for making of
concrete cubes in laboratory," BS-1881-116: Part 110,1990.
[12] British Standards Institution, "Methods for
Determination of compressive strength of concrete
cubes," BS 1881-108: Part 110, 1990.
[13] ASTM C-39, Standard Specification for
Compressive Strength of Cylindrical Concrete
Specimens, Philadelphia, PA: American Society for
Testing and Materials, 2003.

Fig 7. Comparison of Percentage Improvement of
Optimized Sands
ACKNOWLEDGEMENTS
The authors thank Dr. S. Ali Rizwan for his criticalreview which greatly helped in improving the manuscript.
The study was funded by the National University of
Sciences and Technology.

REFERENCES[1] Hudson, B.P, “Concrete Workability with High FinesContent Sands”, 1999.
[2] Golterman, P., Johansen, V., Palbfl, L., “Packing ofAggregates: An Alternative Tool to Determine theOptimal Aggregate Mix,” ACI Materials Journal, Vol.
94, No. 5, 1997.
[3] Shilstone, J. M., “The Aggregate: The
Most
Important Value-Adding Component in Concrete,”
Proceedings, Seventh Annual International Center for
Aggregates Research Symposium, Austin, Texas,
1999.
[4] Washa, G.W., Workability, Chapter 5,
Concrete
Construction Handbook, 1998, 4th edition, New York.
[5] Galloway, J. E. Jr., “Grading, Shape, and Surface
Properties,” ASTM Special Technical Publication No.
169C, Philadelphia, 1994.
[6] ASTM C-136, Standard Test Method for
Sieve
Analysis of Fine and Coarse Aggregates,
Philadelphia,
PA: American Society for Testing and Materials, 2001.
[7] ASTM D-2487, Standard classification of soils forengineering purposes (Unified soil classification
system), Philadelphia, PA: American Society for
Testing and Materials, 1993.
[8] ASTM C-33, Standard Specification for Concrete
Aggregates, Philadelphia, PA: American Society for
Testing and Materials, 2003.
[9] ASTM C-40, Standard test method for organic
impurities in sand for concrete, Philadelphia, PA:
American Society for Testing and Materials, 1992.
[10] ASTM C-192, Standard Practice for Making andCuring Concrete Tests Specimens in the Laboratory,
Philadelphia, PA: American Society for Testing andMaterials, 2000.

611

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Strengthening of Landslide Retaining Pile Wall after Geodetic Monitoring
Frangov G., Zayakova H., Hamova M.,
Higher School of Civil Engineering “L. Karavelov”, Sofia, Bulgaria

ABSTRACT: The investigated landslide occurred in the spring of2004 on the road through the Balkan Mountains, Bulgaria. It waspart of a larger ancient landslide that continues down the slope. To
restore the traffic it was decided not to strengthen the whole landslide, but only the upper part near the road. It was designed andimplemented strength support structure consisting of two rows ofsteel concrete piles incorporated by a beam. To prevent the influenceof groundwater a horizontal drainage above the landslide was carriedout. Geodetic monitoring network was built and periodic measurements were held in the period 2005 - 2010 to monitor the behaviour of the structure. During this period the largest horizontal
displacements were found in the area between first and second
section of the pile construction. Geodetic monitoring was used todetermine the most vulnerable area of the structure. It was reinforced
with additional steel concrete piles combined with a beam.

Keywords: landslide reinforcement, geodetic monitoring
1. INTRODUCTION
Roads often pass through hilly and mountainous areas andthere a greater risk of old landslides activation exists. Some of
the main causes of landslide appearance on mountainous
roads are: steep gradient of slope, river erosion, weak soil
cover above rock basement, usage of an unsuitable soil for
construction of embankments, not enough geological survey,
technological errors during execution of earth work as poor
drainage and compaction of embankments, increasing theload due to more traffic and increase the weight of vehicles.
The most common reasons are complex and lead to landslides,
which reduce the width of the road or completely interrupt
traffic through this section. The road Trojan-Karnare whichcrosses the Balkan Mountains is a typical example [1]. The
aim of this study is to evaluate the trends of the observed
deformations of the pile construction and to strength it.
2. LANDSLIDE ACTIVITY AND CONTROLMEASURES
2.1 Geology and geomorphology
The region refers to highest part of the Balkan mountain. The
southern slope of the mountain is consequent, deeply incisedby many river valleys. The northern slope is steep only in its

upper part, dismembered by river valleys in long transverse
elevations. The territory in a close perimeter from the site isbuilt of differing in genesis and composition rock complexes:
Cherni Ossam Formation (соJ3t – K1bs) is characterized bythin fore-component rhythmic: sandstones, aleurites, marlsand clayey limestones.
Zlataritsa–Cherni Ossam Formation (z-соJ3k – K1bs)
consists of alternation of thick benches with coarse
two-component (flyschoid) rhythmic with benches,characterized by thin three-component rhythmic.
Alluvial formations (аQh) are observed along the Beli and
Cherni Ossam Rivers. Alluvium is represented by unsorted
gravels and differing in grainsize sands and sandy clays.
Deluvial (dQh) and eluvial (eQh) formations are built of
sandy clays and differing in size gravels. They are observed at
the foot of the slope sections. At some places the deluvial
clays have significant thickness. The eluvial deposits are notthick and often they are entirely absent.
A specific feature of the structures is their Early Cretaceous
origin. They are distinguished by their linearity and moreover– by their complexity due to the numerous faults, most often
normal faults, which is typical for the fold-block structure of
the area.
The region is characterized by low water abundance. The
springs are very rare and have insignificant flow rate. They
depend either on tectonic fissures or are formed in the base of
Quaternary deposits. The groundwater is of the fissure typeand is drained in the rivers.
2.2 Landslide features
The great distribution of clayey and flysch sediments, thestrongly rugged terrain, the tectonic structure and thehydrogeological specific features create the conditions for thedevelopment of landslides along the valley slopes [2]. Their
activation is associated with incorrect cutting of slopes andexcessive loading from big embankments. The landslide
activity has been repeatedly displayed, the last activation
being in March, 2004. The developed landslide is a part of abigger older landslide, which continues downwards along theslope. In the upper part its width is 36 m and the depth varies
between 4,5 m and 6,3 m. The landslide body is formed byintensively mixed clays [3].

613

2.3 Landslide control
A radical solution to this landslide would be the constructionof buttresses at the base of the slope foot, with the adequate
drainage of the terrain. Due to different reason like private
forests, the increasing range of the area and depth of thelandslide under the road etc., it has been decided to develop
stabilizing facilities in the close proximity of the roadbed for
ensuring the road stability. The depth of the slide surface, thepassing through a thick layer of asphalt concrete and an
embankment of heterogeneous soil, as well as therequirement to carry out the construction works without
stopping the traffic, made it necessary to perform thestabilization with a structure of two rows of piles, cantileveranchored in the strong marl layer, united by a beam (Fig.1).
The piles were borehole-cast with a diameter of 0.4 m and alength of 10 m. The whole structure is 42.0 m long and [2] it isdivided in three sections with a length of 14.0 m each. The
grid is situated under the road bench and there are drainage
openings in it with diameter 75 mm at a distance of 1.5 mfrom one another.
To reduce the negative effects of ground water on landslide
stability horizontal drainage above the road is built. Its
purpose is to intercept water from the slope and take it awayfrom landslide.

Fig.1. Longitudinal cross section trough the pile wall1 – initial terrain, 2 – terrain after new landslide activation

After implementation of the retaining structure landslide
continued to move down the slope. For this reason the terrainsank and the top of the pilеs under the unifying beam partially
had shown.
3. GEODETIC MONITORING
3.1 Geodetic grid
Measuring and control system was built on the site for
conducting geodetic surveying regime measurement. It
consists of 2 pcs. basic benchmarks located outside the scopeof the landslide and 6 pcs. observed benchmarks on the crownof unifying beam (fig. 2). Nine measurements have been
carried out with a total station TOPCON GTS - 105 N during
the period 2005 - 2010. Direct measurement accuracy forangles is 5" and for length 2 mm. Leveling of observed
benchmarks is done by trigonometric leveling. Coordinate
system is relative, as a starting benchmark was used BB 1 andconditionally accepted elevation 100.000 m.

Fig.2. Location map of the geodetic measurement grid
BB – basic benchmark, OB – observed benchmark

3.2 Discussion
The retaining pilе structure is subjected to the action of landslide pressure and dynamic loads from vehicles after its construction. Because of these actions, it is strained anddeformed. For this reason, relocations have been reported at
the starting moment in all observed benchmarks (fig. 3). The
landslide has a different depth in the area of pile construction
and this affects the landslide pressure. It is the largest in thearea between first and second section where the observed
benchmarks 2 and 3 are located. There the largest displacements are measured.
The speeds of surveying displacements are not constant. For
five years the vector displacement of observed benchmark 2
is 74.46 mm, while for the benchmark 3 is 69.35 mm. The
main part of these movements, respectively 40.01 mm and614

where:k f - coefficient depending on the form of the section of thevector displacement, mm

125,0
OB 1

OB 2

OB 3

OB 4

OB 5

OB 6

100,0
75,0

pile. For a circular section k f =0.9;

k  k1 

50,0
25,0

30.12.2010

30.12.2009

30.12.2008

31.12.2007

30.12.2006

30.12.2005

2(d  1)(6)where:
k1 - coefficient depending on the number of drilling piles in arow parallel to the plane of action of the load, for n>4,
k1 =0.45;

0,030.12.2004(1  k1 ) L ptime

L p - average distance between the piles;

Fig.3. Diagrams of vector displacement of the observed benchmarks

39.20 mm, were appeared during the first year of
measurement. The trend is the gradual decay of thedeformation. The most vulnerable area of the structure is in
the area between first and second section.
The landslide was activated several times in the past. Then a
new layer asphalt is performed to restore the level of road. So
at this zone the thickness of asphalt in the sunken part of theroad reached up to 3.10 m. This asphalt creates a barrier togroundwater and prevent their drainage, regardless of theconstructed drainage. The water pressure on asphalt bodyincreases additionally landslide pressure on the retaining
structure.

The resulting displacements have not reached critical values,
the structure is not threatened by the destruction and the trendof deformation is towards extinction, but to reduce the risk of
new landslides activation and the interruption of the road it
was decided to strengthen the construction in the area
between first and second sections (fig. 4).

4. STRENGTHENING OF EXISTINGCONSTRUCTION
Horizontal displacement of a single pile is obtained according
to following equation [4]:

1 

l0
  nn3EI(1)where:
1 - horizontal displacement of a single pile with a free top
on the level of the pile cap under the horizontal force H = 1,
applied at this level.

 nn 

1
A0
 EI

(2)

Kb
EI

(3)

c  53cwhere:
A0 - coefficient depending on the effective depth of thepile h ;EI - stiffness of the pile; c - elastic characteristics of the pile;h  ch(4)

К- modification of the subgrade reaction;b - effective width of the pile. For boring piles it has to beobtained trough:b  kk f (d  1)

(5)

Fig. 4. Plan of the most vulnerable zone of existing pile wall and additional
strengthening with steel concrete support pile structureThis reinforcement consists of four additional piles withdiameter 320 mm and length 6.0 m, full of concrete B25 andreinforced with double T № 24, unified with reinforced steel

615

concrete beam Among the earlier support structure of thelandslide and new one cross beams and short columns areapplied (fig. 5).

In such cases, very useful is to conduct geodetic monitoring.
The data obtained make it possible to check the status of
construction, the values of the displacements, the trends intheir development. The most vulnerable area of the structure
can be established and additional support to be implemented.
In the case such action is timely taken the construction can beprevented from destruction.
6. ACKNOWLEDGMENT
This study was possible thanks to the courtesy Road Infrastructure Agency of Bulgaria and Regional Traffic Management Lovech. The authors owe their gratitude for theopportunity to participate in solving the landslide problem ofroad Trojan-Karnare.
7. REFERENCES
[1]

[2]

[3]

[4]

Fig.5. Cross section trough previously built pile wall (1) and new support
structure (2)Additional support structure has been completed and it isexpected that the displacements will be reduced completely.
The monitoring of the construction should continue and theadequate measures should be taken, if necessary.
5. CONCLUSION
Landslides stabilization is a complex problem that requires agood understanding of geological, geomorphological andhydrogeological conditions of the particular landslide and theinfluence of various factors on its stability.
The selection of earth retaining structures are often
determined not only by geotechnical arguments but also by
financial, economic and social considerations.
Strengthening only a facility, such as this road, without taking
into account the slope stability as a whole, can cause ruptureof the landslide body, partially crossing the landslide masses
between piles and stripping of the upper part of retaining
structure. Ultimately, static scheme of strengthening changes
and this can lead to deformation of the structure. The retaining structure begins to strain after construction under theaction of landslide pressure. The greatest load on it is in thesection with the deepest slip surface and most intensive
landslide movements. This causes a correspondingly larger
displacements in this zone. Subsequently, the load is redistributed on the rest of the structure, hence deformations in thisarea delay and the landslide pressure is intaken more evenly.
616

Hamova, M., G. Frangov, H. Zayakova. Stabilization of a Landslide
Section Along a Road Passing Through the Stara Planina mt., Bulgaria.
Proc. 13th Danube-European Conference on Geotechnical Engineering,
Ljubljana, Slovenia, 29-31 May 2006, (on CD).
Kolev Ch, Rehabilitation of sliding motorway slopes on deep failure in
Bulgaria. Proc. of the 6th International Conference on Case Histories inGeotechnical Engineering , Arlington, VA, August 11-16, 2008, paper
No. 2.35.
Ivanov, P., N. Dobrev. 2006. Results from laboratory tests of samples
from Fore-Balkan landslides. Proc. National Scient. Tech. Conf. on
Landslides and Erosional Processes in Bulgaria, Prof. M. Drinov, 2006,
155-160.
Dingozov G, Bozhinov B, Pile foundation. Sofia: Technica, 1979, ch.5.

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Response of Tapered Piles under Lateral Harmonic Vibrations
Ahmad Dehghanpoor, Yazd University, Yazd, Iran
Mahmoud Ghazavi, Professor of Civil Engineering, K.N. Toosi University of Technology, Tehran, Iran

ABSTRACT: This paper presents a new method for analysis
tapered piles under lateral harmonic vibration. The behavior oftapered piles is assumed to be as elastic and linear. The soil consistsof some elastic horizontal layers that they are homogeneous,isotropic, and linearly visco–elastic. The pile is divided to somesegments and the differential equation for a given desirable
segment is obtained and solved. Then the dynamic complex
stiffness parameters are derived for the pile head. Parametricstudies have been performed to investigate the influence of pilegeometry, soil properties, and loading details on pile-soil system
amplitudes. It has been found that under lateral harmonic
vibrations, with increasing the pile taper angle, the resonant
amplitude decreases. In addition, it has been concluded that underlateral harmonic vibrations, a tapered pile experiences lower
amplitude than a cylindrical pile of the same length and material
volume.

Keywords: Tapered Piles, Lateral Harmonic Vibration,
Stiffness, Damping
1. INTRODUCTION
There are various methods for studying behavior ofdynamically loaded uniform piles. These are continuum
approach (Novak, [1]; Novak and Aboul – Ella, [2]; andNogami, [3]), boundary element method (Kaynia andKausel, [4]; Sen, Kausel and Banerjee, [5]), lumped–mass
method (Penzien, [6]); and finite element solutions (Blaney,
Kausel and Rosset, [7]; Wolf and Von Arx, [8]; chow, [9]).
Novak presented an approximate continuum approach toaccount for soil-pile interaction: it is assumed that the soil iscomposed of a set of independent horizontal layers of
infinitesimal thickness, which extend to infinity. As eachplane is considered independent, this model may be viewed
as a generalized Winkler model. The planes arehomogeneous, isotropic, and linearly elastic, and they
consider being in a plane strain state. Using Baranov’s [10]solution for the horizontal soil reaction to a rigid circular
disc with harmonic horizontal displacement (representing apile cross section), Novak formulated the differential
equation of the damped pile in horizontal vibration. He
found the steady state (particular) solution for harmonic
motion induced through pile ends, and used this solution tofind the dynamic stiffness of the pile head for different
boundary conditions.
The existent methods are few for analyzing tapered piles
under dynamic loads, so recently it has been attracted many
investigators. The most studies in this piles, related tovertical harmonic vibrations such as finite difference method(Saha and Ghosh, [11]) and mathematical method (Xie andVaziri, [12]).

Investigations have been done on tapered piles includes: Fullfinite element, analytical solution and laboratory tests(Kurian and Moola, [13]), and centrifuge model tests (ElNaggar and sakr, [14]). Field load tests were also conducted
on tapered piles to investigate their load-carrying capacity(Rybnikov,) and indicate that bored cast-in-place tapered
piles can have bearing capacity 20-30% higher than that forcylindrical piles with the same volume and same mean
radius. The Ghazavi has also recently performed full-scale
tests on a tapered pile driven into a cohesive soil profile in
the field. These tests showed that, in long term, the tapered
pile had 80% more capacity than a uniform pile of the same
volume and length. Zil'berberg and Sherstnev [15] have reported from their field tests that driven tapered piles in sandy
soils can give a stiffer and stronger axial response resulting
in a 200-250% increase in bearing capacity when comparedto the capacity of cylindrical piles with the same volume andmean radius. The response of these piles under lateral static
loads was also investigated El Naggar and Wei, [16]. El
Naggar and Wei [17] also conducted tests in a pressurechamber on tapered model piles subjected to uplift loads.
Ghazavi et al. [18] described the performance of tapered Pile
during Pile driving and Ghazavi and Tavassoli [19] Study ofPile geometry on Pile driving. Also has proceeded analysisof kinematic seismic response of tapered piles (Ghazavi,[20]) and Response of tapered piles to axial harmonic loads
and effect of angles on tapered piles has been discussed thenobtained has been verified whit finite element methods andthey were satisfactory (Ghazavi, [21]).
SSM (Segment by Segment Method) is based on continuum
method of Novak elasto- dynamic approach for analysis ofpiles. The SSM has been applied to uniform piles under axial
compressive loads (Ghazavi et al. [22]), uplift static loads(Ghazavi et al. [23]), and axial and lateral harmonic
vibrations (Ghazavi, [24], Ghazavi and Dehghanpour [25]) in
this paper SSM was applied to analyze dynamic behavior oftapered piles under lateral harmonic loads.

2. ANALYTICAL MODEL
The characteristic effects of surrounding soil on the pile
response are determined with stiffnessand damping
parameters of soil – pile system. These effects can be taken
in to account if a proper soil reaction is employed.
For analyzing tapered pile, it was idealized to somecylindrical segments with different diameter that connected
together by rigidly at nodes. This idealization was used in
tapered piles under harmonic axial vibration [11-12] and
[21].

617

Surrounding soil reaction to the loaded tapered pilesegments had been presented by Vs(z, t) and Vr(z, t), thatshear resistance per unit length of the pile shaft andhorizontal reaction at the horizontal annular projections of
the pile shaft, respectively. Parameters z and t represent
depth and time in order as shown in Fig l.

using other, existing solutions. This assumption has also
been used by other researcher [2]. In one dimensional finite
element analysis of cylindrical piles under torsional
vibrations, Novak and Howell also used the same
assumption. A somewhat similar assumption was also madeand by Mylonakis and Gazetas [30] for axially loaded
cylindrical piles in a layered soil profile. It is note that
Novak and Aboul – Ella used finite element method (FE) foranalysis of piles embedded in inhomogeneous soil andsubject to lateral harmonic vibrations. By considering thetypical embedded segment j at depths, shown in Fig 1 and on
the basis of the above assumption, the following governing
dynamic differential equation of a soil – pile system
subjected to harmonic lateral load can be obtained [1]:m pj 2 u j ( z, t )t2

 C pj 2 u j ( z, t )t E pj I pj

 4 u j ( z, t )z 4

 Gsj S uj u j ( z, t )  0(2)
Where mpj the pile mass per unit length, Cpj is the damping
coefficient of the pile material, EpjIpj is the bending stiffnessof the pile segment j, Gsj is the shear modulus of the soil
layer surrounding the pile segment j , uj (z, t) is the local
time – dependent complex amplitude at depth z from the topof segment j and Suj is Complex dimensionless soil
resistance parameter defined elsewhere [1] as a function ofr Poisson's ratio and dimensionless frequency, a  j  oj V

Fig.1 Idealization of tapered pile for dynamic analysis inlaterally inhomogeneous media using segment by segment, (a)Actual pile; (b) Idealized pilesj

The soil response with time to motion of the pile toe, Fb(t), istaken as that of a viscoelastic half – space to rigid, mass less,circular disc of radius rb undergoing harmonic vibration. Thiscan be expressed as:
Fb (t )  Gb rb [C u1 (a b , v, D)  iC u 2 (a b , v, D)]u b (t )(1)
Where Gb is the soil shear modulus at the pile toe, Cul , Cu2
are dimensionless complex parameters given in the form ofpolynomial expressions (Veletsos and Vbric, [26]), ub(t) is
the toe horizontal displacement, rb is pile radius at the piler
tip, ab  b , where Vb is shear wave Velocity of soil

Here roj is the pile segment radius,  is the circularfrequency, and Vsj the shear Wave velocity of the soil
surrounding the pile segment j.
The four terms in (2) represent the inertia force due tolumped mass of the pile, the damping force of pile material,
the lateral interaction between pile segments, and the soil
resistance, respectively. For harmonic vibration, the local
displacement uj (z, t) is given by:it(3)u j ( z, t )  u j ( z ) e
Where uj(z) is the complex amplitude at depth z from the topof segment j and  is the excitation frequency:
u j ( z )  u1 j ( z )  iu 2 j ( z )Vbbelow the tip,  is circular frequency,  is Poisson's ratio,
D is material damping.
The soil reaction on the pile is represented by springs anddashpots, which are modeled on elasto – dynamic theory.
The interaction of the soil and the pile is then determined foreach segment according to the characteristics of soil. This
interaction can be demonstrated by a complex displacement,
shear force, rotation and bending moment at the end of theadjacent segment. This procedure is performed from thelowest pile segment and extends to the next upper segment.
This manner is continued to reach the topmost segment. That
is why this procedure is called the SSM (Ghazavi [18-25]).
In the analysis, it is assumed that the soil reaction associatedwith a given soil layer is identical to that of an infinite rigid
pile undergoing a uniform displacement of the same
properties as the soil of that layer. This assumption isessential to the solution and will be examined subsequently(4)

Combining (2) and (3) gives
E pj I pj 4u j ( z)z 4

 u j ( z )[G sj S u1  m pj  2  i (C pj   G sj S u 2 )]  0(5)
The above equation can be solved explicitly. The solution forthe displacement at a point at vertical distance z below theupper node of segment j is given by:
u j ( z)  Aj Cosh( jzzzz)  B j sinh( j )  C j cos( j )  D j sin( j )
hj
hj
hj
hj

(6)
Where

618

 j  hj 4

1
[m pj  2  G si S uq  i (C pj   G sj S u 2 )]E pj I Pj

u j1 

(7)
Where Aj, Bj, Cj and Dj are integration constants determinedusing appropriate boundary conditions.
If the displacement, rotation, shear force and bending
moment transmitted by the pile at node 2 of segment j areknow, the integration constants Aj, Bj, Cj and Dj can becalculated. Thus, the displacement, rotation, bending
moment and shear force at node 1 of segment j arerespectively given by (8a-8d).

 j1 

Vj 2h3j (sinh j  sin j )   2j Ep j I p j  j 2hj (sinh j  sin j )   3j Ep j I p j u j 2 (cosh j  cos j )  M j 2h2j  j (cos j  cosh j )2Ep j I p j  3j

Vj 2h3j (cos j  cosh j )  2j Ep j I p j  j 2hj (cosh j  cos j )   3j Ep j I p j u j 2 (sin j  sinh j )  M j 2h2j  j (sinh j  sin j )2Ep j I p  2j hj

M j1 

Vj1 

Vj 2h3j (sinh j  sin j )   2j Ep j I p j  j 2hj (sinh j  sin j )   3j Ep j I p j u j 2 (cosh j  cos j )  M j 2h2j  j (cosh j  cos j ) 2 j h2jVj 2h3j (cosh j  cos j )  2j Ep j I p j  j 2hj (cosh j  cos j )  3j Ep I p u j 2 (sinh j  sin j )  M j 2h2j  j (sinh j  sin j ) 2h3j

3. PARAMETRIC STUDIES FOR PILE GEOMETRY(8a)- (8d)

EFFECT
In this section, four type of piles whit difference geometry
under lateral harmonic load and have same length andvolume had been studied and the results are compared.
Properties of tapered piles and soil have presented in Table
1. All piles are 10m of length and there volumes are 1.36 m3.
Pile C is cylindrical. Pile T-C consists of a top tapered
segment with 5m length and a lower cylindrical segmentwith 5m length. Pile C-T has top cylindrical part with 5m
followed by a tapered part with 5m length. Pile T is tapered.
Taper angles of piles are 0.5 and 1.5 that has been shownin Fig 2 and Table 2.

Table1. Properties of tapered piles and soil
Radius of equivalent circular pile, req*

0.208 mShear wave velocity in soil, Vs

84m/s

Soil Poisson’s ratio, s

0.45

Soil unit weight, s

17.5 kN/m3

Pile modulus of elasticity, E p

1.962×107
kN/m2

Soil modulus of elasticity, E s

3.58×104
kN/m2

Pile unit weight, p

25 kN/m3(* req is radius of cylindrical pile of the same volume and length as
tapered pile)Fig. 2 Pile configurations

Table2. Dimension of tapered piles for Fig 3.

Piles  0.5

Taper
Angle
T
C

L=10m
L=10m

TC

L1=5
m
L1=5
m

CT

D1=0.5m
D1=0.416m
L2=5
m
L2=5
m

  1.5
D2=0.326m
D2=0.416m

D1=0.48m

D2=0.393m

D1=0.4369m

D2=0.3497
m

619

L=10m
L=10m
L1=5
m
L1=5
m

D1=0.65m
D2=0.126m
D1=0.416
D2=0.416mm
L2=5
D1=0.603m D2=0.342m
m
L2=5
D1=0.4729
D2=0.2111
m
m
m

SSM was applied with assumption pinned–ended for all ofabove piles and Fig. 2 illustrate the dimensionless amplitude
lateral versus the excitation frequency for piles with taper
angles   0.5 that results have been shown in Fig. 3 the T
pile have the least lateral and rotary response amplitude andafter T Pile. There are T –C, C-T and C respectively.
T
TC

14

C

Amplitude, A(u)

12

CT

10
8
6

CT and C piles respectively have the least lateral and rotary
response amplitude and observed that for tapered piles of thesame volume and length under lateral harmonic vibrations,with increasing the taper angle, the resonant frequency
increases slightly. However, the reduction of the amplitude
is more pronounced.
The SSM is an efficient and simple method for analysis oftapered piles under harmonic vibration. In particular, theeffects of the soil in homogeneity in the vertical direction
even with complicated stratifications can be easily captured.
This method involves less computational work thanavailable numerical method based on the FE.
6. REFERENCES

4[1] Novak, M. (1974): Dynamic stiffness and damping of piles, Canadian
Geotechnical
Journal,
Vol.
11,
No.
4,
574-598.

2
0
0

5

10

15

20

25

30

Frequency (Hz)[2] Novak, M. and Aboul-Ella, F. (1978): Impedance functions of piles inlayered media, Journal of Geotechnical Engineering Division ASCE, Vol.
104, No. 6, 643-661.

Fig. 3 Variation of lateral dimensionless amplitude versus[3] Novak, M. and Nogami, T. (1980): Coefficients of soil reaction to pile
vibration, Journal of Geotechnical Engineering Division ASCE, Vol. 106, No.
5, 565-570.

frequency for T pile (   0.5 )It is noted that in the C piles, the resulted values of are exactly
the same as those reported by Novak [1].
In Fig. 4, the dimensionless amplitude lateral versus theexcitation frequency had been compared based on taper
angle in T pile and observed with increasing taper pile,dimensionless response amplitudes will decrease.

12  1 .5 

Amplitude, A(u)10  0 .5

[7] Blaney, G.W., Kausel, E., and Roesset, J.M. (1976): Dynamic stiffnessof piles, Proceedings 2nd International Conference on Numerical Methods
in Geomechanics, Blacksburg, Virginia, 1001-1012

  0

4[8] Wolf, J.P. and von Arx, G.A. (1982): Horizontally travelling waves in agroup of piles taking pile-soil-pile interaction into account, International
Journal of Earthquake Engineering and Structural Dynamics, Vol. 10, 225-237.

2
0
0

5

10

15

20

25

[5] Sen, R., Kausel, E., and Banerjee, P.K. (1985): Dynamic analysis of pilegroups embedded in non-homogeneous soils, International Journal for
Numerical and Analytical Methods in Geomechanics, Vol. 9, No. 6, 507524.
[6] Penzien, J. (1975): Soil-pile interaction. Earthquake Engineering, R.L.
Wiegel, ed., Prentice Hall Inc., Englewood Cliffs, N.J. 349-381.



86[4] Kaynia, A.M. and Kausel, E. (1982): Dynamic behaviour of pile groups,
Proceedings 2nd International Conference on Numerical Methods in
Offshore Piling, Austin, Texas, 509-532.

30

Frequency (Hz)Fig. 4. Comparison variation of lateral dimensionless amplitude
versus frequency for T pile according to taper angle

5. CONCLUSION
A simple approach, called SSM, has been presented in thispaper for determination of stiffness and damping parametersof laterally loaded tapered piles subjected to harmonic
vibrations. The soil-pile interaction in this method ismodeled within each segment and applied via the segment
nodes to the analysis of the adjacent segment. Therefore, thestiffness and damping parameters for the whole pile-soil
system are determined. According to results SSM, T, TC,[9] Chow,Y.K. (1985): Dynamic response of axially loaded piled foundations,
Proceedings 5th International Conference on Numerical
Methods inGeomechanics, Nagoya, Japan, 1485-1492.
[10] Baranov, V.A., (1967): On the calculation of an embedded foundation,(in Russian), Voprosy Dinamiki Prochnosti, Polytechnical Institute of Riga,
No. 14, 195-209[11] Saha, S. and Ghosh, D.P. (1986): Vertical vibration of tapered piles,
Journal of Geotechnical Engineering, ASCE, vol. 112, No. 3, 290-301.
[12] Xie, J., and Vaziri, H.H. (1991): Vertical vibration of nonuniform piles,
Journal of Geotechnical Engineering ASCE, Vol. 117, No. 5, 1105-1118.
[13] Kurian, N.P. and Srivinas, M.S. (1995): Studies on the behavior of
axially loaded
tapered piles by the finite element method, International
Journal of Numerical and Analytical Methods in Geomechanics, Vol. 19,
869-888.

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[14] El Naggar, H. and Sakr, M. (2000): Evaluation of axial performance oftapered piles from centrifuge tests, Canadian Geotechnical Journal, Vol, 37,
1295-1308.
[15] Zil'berberg, S. D., and Sherstnev, A. D. (1990):Construction ofcompaction tapered pile foundations (from the experience of the"Valdespetsstroi" Trust), Soil Mechanic and Foundation Engineering, 27(3),96-101.
[16] El Naggar, H. and Wei, J.Q. (1999): Response of tapered piles
subjected to lateral loading, Canadian Geotechnical Journal, Vol, 36, 52-71.
[17] El Naggar, H. and Wei, J.Q. (1998): Experimental study of axial
behaviour of tapered piles, Canadian Geotechnical Journal, Vol, 35, 641654.
[18] Ghazavi, M., Williams, D. J., and Wong, K.Y. (1996): Analysis of atapered pile during pile driving, Proceedings 2nd International Conference
on Multi - Purpose High-Rise Towers and Tall Buildings, Singapore, 87-94.
[19] Ghazavi, M. and Tavassoli, O., (2008): Numerical Analysis for Pile
Geometry Effect on Drivability. Proceedings of the 8th International
Conference on the Application of Stress Wave Theory to Piles, Lisbon,
Portugal, pp. 271- 276.
[20] Ghazavi, M., (2007): Analysis of Kinematic Seismic Response ofTapered Piles. Geotechnical and Geological Engineering, an International
Journal, vol. 25, No. 1, 37-44.
[21] Ghazavi, M., Response of Tapered Piles to Axial Harmonic Loading.
Canadian Geotechnical Journal, in press[22] Ghazavi, M., Williams, D. J., and Morris, P.H. (1997a): Analysis of
axially loaded piles embedded in layered deposits, Proceedings 2nd
International Conference on the Application of Numerical Methods inEngineering, Universiti Pertanian Malaysia, Selangor, Malaysia, 335-344.
[23] Ghazavi, M., Wiiliams, D.J., and Morris, P.H. (1997b): Analysis of piles
subjected to uplift loads, Proceedings 2nd International Symposium on
Structures and Foundations in Civil Engineering, Hong Kong, 177-183.
[24] Ghazavi, M. (2002): Vertical vibration of deep foundations in layered
deposits, Proceedings of Iranian Third International Seminar on Soil
Mechanics and Geotechnical Engineering (ISSMGE), Tehran, Iran, Vol. I,
223-228.
[25] Ghazavi ,M., Dehghanpour, A., “Dynamic Analysis of Piles under
Lateral Harmonic Vibration”. Fifth International Conference on Recent
Advances in Geotechnical Earthquake Engineering and Soil Dynamics,
May 24-29, 2010; San Diego, California, USA.

[26] Veletsos, A.S., and Verbic, B., (1973): Vibration of Viscoelastic
Foundations. Earthquake Engineering and Structural Dynamics, Vol 2, No.
1, 87-102621

First International Conference on Geotechnique, Construction Materials andEnvironment, Mie, Japan, Nov.21-23, 2011, ISBN: 978-4-9905958-0-7 C3051

Investigation of Ground Movement in Soil-nailed Walls Using 1g Models andPIV Method
H. Mahdavi and M. Hajialilue-Bonab, Civil Engineering Department, University of Tabriz



ABSTRACT: Soil nailing due to its unique advantages has gained
great popularity in recent years. In urban areas, the impact of thecreated excavation on the surrounding structures and utilities is ofgreat consequence. This paper is mainly focused on the investigationof the ground movements in soil-nailed walls. 1g models wereconstructed to study the impact of nail inclination on the movements
induced by the excavation. Despite the limited use of 1g models
owing to some similarity problems, they can provide valuable resultswith low costs. An intermediate method was used to simulate the
excavation process and images were captured during this process.
Then the captured images were analyzed using particle image
velocimetry (PIV) method. The results provide a better
understanding of the deformations in the soil-nailed walls.
Keywords: nail inclination, physical modeling, PIV, soil
nailing

Experimental studies conducted by Marchal (1984) [5]showed that the direction of the nails with respect to thepotential failure surface plays a role in the mobilization oftension and shear. Jewell (1980) [6], [1] experimentally
verified that the optimum direction for the mobilization oftension in the flexible nails is approximately 30˚ in relation to
the normal at the shear surface. The above mentioned list of
the studies just gives a brief account of the conducted
research and is not complete for the sake of succinctness. This
paper is focused on the investigation of the effect of the nailinclination on the movements of the soil behind thesoil-nailed wall. 1g models were constructed to study thisissue and particle image velocimetry (PIV) method was
employed to further investigate its behavior. PIV is an image
analysis technique developed by D. J. White and W.A. Takein Cambridge [7]1. INTRODUCTION
In recent years, due to an increase in urban construction there
seems a necessity to create deep excavations to build new
structures. Experience shows when the importance of the
impact of the ground movements resulted from the createdexcavation on the adjacent structures is underestimated, greatcosts have been inflicted upon engineers. Soil nailing, due to
its rapid construction and competitive costs, has gained great
popularity in recent years. Ground movements induced bysoil nailed excavations are relatively large compared to otherretaining structures; however, this could be somewhat
controlled using the optimum layout of the nails. This
research was done to find the optimum layout for the nails. As
discussed in Clouterre (1993) [1], the nails are more efficientat controlling lateral displacements of the structure when
there is no inclination. A numerical study on the optimum
layout of soil-nailed slopes was conducted by Fan and Luo(2008) [2]. Effect of nail orientation and geometric layout on
the overall stability of soil–nailed slopes with various
geometric conditions was investigated using a non-linear
finite element approach. Stability of soil–nailed slopes wasevaluated in terms of factor of safety. They found thehorizontal placement of the nails to be the optimal nailorientation for slope angle of 90˚. The effect of the nail skew
angle of soil nailing wall was investigated by Bang andChung (1999) [3]. They developed a finite element method ofanalysis utilizing the generalized plane strain approach toanalyze three dimensional stress, strains, and displacements.
Shafiee (1986) [4] used finite element method to investigate
the impact of the nail inclination on soil-nailed wall
deformation. He concluded that the horizontal placement of
the nail leads to a smaller deformation than the inclined nails.

2. MODEL PREPARATION
The dimensions of the box used for the tests were100ൈ60ൈ30cm. It was made of steel plates with the thicknessof 3mm and stiffeners were used to minimize its deformation
under the applied stresses. The box had a side made ofplexiglass with a thickness of 3cm so that the model
deformation could be monitored during the tests. Anotherplate of plexiglass was glued onto the opposite side of theplexiglass side in order to create symmetrical models and to
reduce the sidewall friction. Dry sand was used to construct
the models. The properties of sand are presented in Table Ι.
The dry sand was pluviated using a special perforated
container to construct uniform and repeatable models. Nails
were made of rods with the diameter of 3mm.The end part ofthese rods were threaded using a threading die in order to use
nuts to fix the nails on the facing. A uniform layer of sand was
glued onto the nails to create rough nails which increased thediameter to 4mm. Hard facing was used in this research. It
was made of an aluminum plate with the thickness of 1.5mm,
and width and height of 30cm each. Since there is a goodconnection between the shotcrete and the facing in prototype
walls, another uniform layer of sand was glued onto the innersurface of this plate and a rough surface was created. The
thickness of the facing increased to 2mm after the sand was
glued. The layout of nails is shown in Fig.1. Thirteen nails
were used in the models.

623

Φ
28˚

Table Ι. Properties of the used sand
CC
CU
γdmin
γdmax
GS
16.76
2.637
0.87
1.36
14.20
KN/m3

KN/m3

Fig. 1. The layout of the nails and the dimensions of the facing

3. TESTING PROCEDURE

Fig. 3. Excavation of soil in front of the soil-nailed wall

The first stage of the model construction involved pluviating
the dry sand from a special perforated container, which washeld approximately 50cm above the test box to gain
uniformity. When an appropriate height was achieved, thefacing was placed in the box. Then the pluviating processcontinued in both sides of the facing until the first row of thenails was reached. Then the nails were placed with theintended inclinations. In the next step, sand pouring was
resumed on both sides of the facing and the model
construction was completed. In order to model theconstruction process of soil-nailed wall, the fill on the front
side of the facing was excavated to model the top down
construction phases of the wall. The excavation was
conducted in 8 stages using vacuum in the vicinity of the wall
to prevent any disturbance. Images were captured using a
Canon PowerShot G10 camera with the resolution of 14.1
megapixels and a CCD sensor. Remote Capture software anda PC were used to capture the images in order not to disturb
the camera during the test. Two 1000W lamps were used toilluminate the models during the production of images. Figs.
2 and 3 show how the excavation process is modeled.

4. RESULTS OF THE 1G TESTSThe details for the conducted tests and the schematic
diagrams for them are presented in Table ΙΙ. Five models wereconstructed using different nail inclinations and all the othervariables (such as Sv, Sh, nail length, back slope…) wereconsidered to be control variables.

Fig. 2. A completed model

624

Table ΙΙ. Details of the tests
Test
No.

Nail
Inclination

Nail
Length

1

0˚

30 cm

2

10˚

30 cm

3

20˚

30 cm

4

30˚

30 cm

5

40˚

30 cm

Test Configuration

GeoPIV [8], [9] and Matlab software programs were used toanalyze the captured images and to calculate the displacement
vectors and shear strains in the soil behind the soil-nailed wall.
In Figs. 4-7, both x and y axes are normalized to the wall௑
height (H), and the origin is assumed to be the wall toe ( =0,ு௒=0). As shown in Figs. 4-7, the displacement vectors
ு
increase with increasing nail inclination. The reason for this
could be explained in terms of tensile force in the nails. As theinclination of the nails increases, the tensile force in the nails
decreases; therefore, the overall shear strength of the slope
decreases [6]. The displacement observed in the test with thenail inclination of 40˚ was large and caused wild vectors in
the results. The comparison of displacements for different
cases is presented in Fig. 8. The displacement of wall top ischosen for this comparison.
Fig. 6. Displacement vectors for nail inclination of 20˚Fig. 4. Displacement vectors for nail inclination of 0˚
Fig. 7. Displacement vectors for nail inclination of 30˚

Top Displacement (mm)

12
10
8
6
4
2
0
0

10

20

30

40

Nail Inclination (degree)
Fig. 5. Displacement vectors for nail inclination of 10˚

Fig. 8. Wall top displacements for different nail inclinations

625

In Figs. 9-12, the maps for shear strains are presented. Strain
maps give a better view of the shear bands formed in the soil
mass. The red color denotes large deformations and the blue
color is assigned to zero strains. Same strain ranges (0-0.5)
were chosen for all the diagrams in order to make acomparison between different tests. It can be clearly seen thatwith the inclination of the nails increasing, the movement of
the soil increases i.e. the red colored area gets bigger.

Fig. 11. Shear strains for nail inclination of 20˚Fig. 9. Shear strains for nail inclination of 0˚

Fig. 12. Shear strains for nail inclination of 30˚Fig. 10. Shear strains for nail inclination of 10˚

The other important issue is the interesting pattern of multiple
shear bands formed in the nailed mass. Using Matlab Image
Processing Toolbox and Curve Fitting Toolbox the shape ofthese shear bands were determined as parabolic. In Figs.
13-16, the parabolic shear band and its formula are presented.
It is worth mentioning that the axes and the presented formula
are in pixels.

626

Fig. 13. Parabolic shape of the shear band, nail inclination=0˚

Fig. 16. Parabolic shape of the shear band, nail inclination=30˚

5. CONCLUSIONS
A number of conclusions can be drawn from the 1g tests
conducted in this research. The results are summarized asfollows,- Displacement vectors increase with increasing nailinclination- Shear strains in the retained soil behind the wall
increase with increasing nail inclination- The shear bands formed behind the nailed-wallwith parabolic shape
Fig. 14. Parabolic shape of the shear band, nail inclination=10˚

6. REFERENCES
[1]
[2]
[3]
[4]

[5]

[6]
[7]
[8]

Fig. 15. Parabolic shape of the shear band, nail inclination=20˚
[9]

627

CLOUTERRE, “Recommandations CLOUTERRE 1991 - Soil Nailing
Recommendations 1991,” English Translation, 1993.
C-C Fan and J-H. Luo, "Numerical study on the optimum layout ofsoil-nailed slopes," Computers and Geotechnics 35, 2008, pp. 585-599.
S.C. Bang and Y.I. Chung, "The effect of the Nail Skew Angle of Soil
Nailing Wall," KSCE Journal of Civil Engineering, Vol. 3, No. 1, 1999,
pp. 73-79.
Shafiee S., "Simulation nume´rique du comportement des sols cloue´s.
Interaction sol reinforcement et comportementde l’ouvrage," The`se deDoctorat de l’Ecole Nationale des Ponts et Chausse´es, Paris, France;
1986.[In French]
Marchal J., "Reinforcement des sols par clouage. E´ tude
expe´rimentale en laboratoire," In Proceedings of International
Conference In Situ Soil and Rock Reinforcement, Paris; 1984, pp.
275–8. [In French]
Jewell RA., "Some effects of reinforcement on the mechanical behaviorof soils," Doctor of Philosophy Thesis, Cambridge University, 1980.
White, D. J., Take, W. A., and Bolton, M. D., “Soil deformation
measurement using particle image velocimetry (PIV) andphotogrammetry,” Geotechnique 53, No. 7, 2003, pp. 619–631.
White, D. J., Randolph, M., and Thompson, B., “An image-based
deformation measurement system for the geotechnical centrifuge,” Int.
J. Phys. Modell. Geotech. 3, 2005, pp. 1–12.
White, D. J., and Take, W. A.. “GeoPIV: Particle Image Velocimetry(PIV) software for use in geotechnical testing,” Rep. No.
D-SOILS-TR322, Cambridge Univ, 2002.

Author Index

Abdul Karim
Ahmad Dehghanpoor
Ahmed M. Elsharief
Frangov G.
Gauhar Sabih
Hajialilue-Bonab M.
Hamova M.
Hau B.C.H.
Hossain M.Z.
Leung F.T.Y.
Mahdavi H.
Mahmoud Ghazavi
Moghaddas Tafreshi S.N.
Muhammad Awais Shafique
Muhammad Ibn Ibrahimy
Muhammad Jamil
Mukabi J.N.
Rajaraman J.
Sharifi P.
Tanvir Iqbal Qayyum
Tham L.G.
Thiruvenkatasamy K.
Yan W.M.
Zayakova H.
Zein M.

577
617
577
613
607
623
613
589
563
589
623
617
601
583
573
607
563
597
601
583
589
597
589
613
577

649

ISBN978-4-9905-9580-7
C3051 \21000E
9784990595807
定価 （本体２１０００＋税）
1923051210008

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