ISSN 0386-5878
Technical Reportof PWRI No.4172

PROCEEDINGS OF
THE 25th U.S. – JAPAN
BRIGDE ENGINEERING WORKSHOP
Tsukuba, Japan
October 19, 20 and 21, 2009

CENTER FOR ADVANCED ENGINEERING STRUCTURAL ASSESSMENT AND RESEARCHPUBLIC WORKS RESEARCH INSTITUTE
Incorporated Administrative Agency

1-6, Minamihara, Tsukuba-shi, Ibaraki-ken, 305-8516 Japan

Opinions, findings, conclusions and recommendations expressed in the papers contained inthis publication are those of the author(s).

Copyright © (2010) by P.W.R.I.
All rights reserved.

No part of this book may be reproduced by anymeans, nor transmitted, nor translated into a machine language without
the written permission of the Chief Executive of P.W.R.I.

この報告書は、独立行政法人土木研究所理事長の承認を得て刊行 したものである。したがって、本報告書の全部又は一部の転載、複製は、
独立行政法人土木研究所理事長の文書による承認を得ずしてこれを 行ってはならない。

Technical Reportof PWRI No.4172

PROCEEDINGS OF
THE 25th U.S. – JAPAN
BRIGDE ENGINEERING WORKSHOP

By
Editor: Masahiro Shirato

Synopsis:
The proceeding documents the results of the 25th U.S. – Japan Bridge Engineering Workshop
which was held at Tsukuba, Japan, on October 19, 20, and 21, 2009, as a part of the activities of
the Panel on Wind and Seismic Effects, UJNR. The Workshop was organized by Task Committee
G “Transportation System” (U.S. side chair: Dr. W. Philip Yen, FHWA, Japan side chair: Mr.
Atsushi Yoshioka, CAESAR, PWRI) of the Panel.

Key Words: Bridge, Design, Construction, Maintenance, Seismic, UJNRPage Left Blank

PREFACEThe 25th US-Japan Bridge Engineering Workshop is a continuation of a series oftechnical interchanges between the United States and Japan on all topics related tobridge engineering. This series of workshops has been conducted under the auspicesof Task Committee “G” of the US-Japan Panel on Wind and Seismic Effects, which isone of the 18 panels making up the United States-Japan Cooperative Program in
Natural Resources (UJNR). The previous workshops are indicated below.
1stFebruary 20-23, 1984

Tsukuba Science City, Japan

2nd

August 19-20, 1985

San Francisco, California, United States

3rd

May 8-9, 1987

Tsukuba Science City, Japan

4th

May 11-12, 1988

San Diego, California, United States

5th

May 9-10, 1989

Tsukuba Science City, Japan6th

May 7-8, 1990

Lake Tahoe, Nevada, United States

7th

May 8-9, 1991

Tsukuba Science City, Japan

8th

May 11-12, 1992

Chicago, Illinois, United States

9th

May 10-11, 1993

Tsukuba Science City, Japan

10th

May 10-11, 1994

Lake Tahoe, Nevada, United States

11th

May 30-31, 1995

Tsukuba Science City, Japan

12th

October 29-30, 1996

Buffalo, New York, United States

13th

October 2-3, 1997

Tsukuba Science City, Japan

14th

November 3-4, 1998

Pittsburgh, Pennsylvania, United States

15th

November 9-10, 1999

Tsukuba Science City, Japan

16th

October 2-4, 2000

South Lake Tahoe, Nevada, United States

17thNovember 12-14, 2001

Tsukuba Science City, Japan

18th

October 22-24, 2002

St. Louis, Missouri, United States

19th

October 27-29, 2003

Tsukuba Science City, Japan

20th

October 4-6, 2004

Washington, DC, United States

21st

October 3-5, 2005

Tsukuba Science City, Japan

22nd

October 23-25, 2006

Seattle, Washington, United States

23rd

November 5-7, 2007

Tsukuba Science City, Japan

24th

September 22-27, 2008

Minneapolis, Minnesota, United States

25th

October 19-21, 2009

Tsukuba Science City, Japan

The steering committee for the 25th US-Japan Bridge Engineering Workshop
consisted of Atsushi Yoshioka, Phillip Yen, David Sanders, and Masahiro Shirato.
The workshop was held at the NILIM and the CAESAR, PWRI, in Tsukuba, Japan.
The 2-1/2 day workshop focused on: 1) Earthquake Case Histories, 2) Inspection andManagement, 3) Accelerated Bridge Construction, 4) Remedial Work & Partial
Replacement, 5) Maintenance 6) Seismic Performance Evaluation, and 7) Seismic
Retrofit. Thirteen participants from the US and twenty-eight participants fromJapan attended the workshop who were arranged by both T/C chairs in terms of the
focused themes. The papers contained within this proceeding are the papers thatwere presented at the workshop (17 papers from the Japan side and 12 papers fromthe US side).
In addition to the workshop, there was a bridge study held after the workshop,
October 22-23, 2009, visiting bridge sites:-

Acelerated bridge construction, Ton-ya machi viaduct-

Seismic Retrofit, Tsuboyama viaduct and Miwatari bridge-

Construction of integral abutment jointless bridge, Okegawa Bridge-

Construction site of corrugated-steel web prestressed concrete brige, Ura Takao
Bridge-Construction site of narrow-two-box-girder bridge, Minami Asakawa Bridge,-

Maintenance for Kisogawa Oh-hashi Bridgein the Kanto and Chubu regions.
Editor: Masahiro Shirato, PhD, Senior Researcher, Bridges and Structures ResearchGroup, Center for Advanced Engineering Structural Assessment and Research(CAESAR), Public Works Research Institute (PWRI)TABLE OF CONTENTS
25 US-Japan Bridge Engineering Workshop
Tsukuba, Ibaragi October 19 to 21, 2009th

PREFACETABLE OF CONTENTS
WORKSHOP PROGRAM

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・1

SESSION 1: Earthquake Case Histories
1-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・9

Seismic Performance Evaluation of Seismic Retrofitted Steel Arch Bridge That
Affected by 2007 Niigata-ken Chuetsu-oki Earthquake

・・・・・・・・・・・・・・・・・・・・11

By Junichi Sakai and Shigeki Unjoh
1-2

Analyses of Damaged Bridge by Ground Displacement During Niigata-ken
Chuetsu-oki Earthquake

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・19

By Takao Okada and Shigeki Unjoh
1-3

Torsional Effects on Seismic Performance of Square vs. Circular RC Bridge
Columns

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・27

By DJ Belarbi and S. Suriya Prakash
1-4

Experimental Study on the Seismic Response of Bridge Columns Using
E-Defense

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・41

By Kazuhiko Kawashima, Tomohiro Sasaki, and Koichi Kajiwara
1-5

Evaluation of the Seismic Performance of Bridge Reinforced Concrete Columns
Under Combined Actions Using Shake Table

・・・・・・・・・・・・・・・・・・・・・・・・・・・・55

By David Sanders and Juan G. Arias-Acosta

SESSION 2: Inspection and Management

2-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・65

Analysis of Periodic Inspection Results of Highway Bridges in Japan
By Koichi Ikuta, Takashi Tamakoshi, and Masanori Okuboi・・・・・・・・67

2-2

Bridge Inspection and Management in California

・・・・・・・・・・・・・・・・・・・・・・・・79

By Barton Newton
2-3

Measures for Strategic Preventive Bridge Management of Tokyo Metropolitan
Government

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・87

By Taro Awamoto and Sentaro Takagi
2-4Instrumentation and Monitoring of I35W St. > Anthony Falls Bridge

・・・・・・・・・97

By Catherine French, Carol Shield, Henryk Stolarski, Brock Hedegaard,
and Ben Jilk
2-5

Study for Bridge Renewal and Repair by Osaka Municipal Government

・・・・・111

By Shinsuke Yumoto, Tetsuya Yokota, and Yasutomo Komatsu

SESSION 3: Accelerated Bridge Construction
3-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・119

Seismic Performance and Structural Details of Precast Segmental Concrete Bridge
Columns

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・121

By Jun-ichi Hoshikuma, Shigeki Unjoh, and Junichi Sakai
3-2

Accelerated Constructions of the Viaducts on the Second Keihan
Expressway

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・131

By Yoshihiko Taira, Hirotsugu Mizuno, Kei Muroda, and Akio Kasuga
3-3

New FHWA Seismic Hazard Mitigation Studies for Highway Bridges

・・・・・・・141

By W. Phillip Yen
3-4

Overview of the Development Process and Effects of the UFO Method

・・・・・・147

By Yuji Mishima, Yukio Katsuta, and Tomoaki Tsuji

SESSION 4: Remedial Work & Partial Replacement
4-1

・・・・・・・・・・・・・・・・・・・・・・・・・161

Experimental Study on the Time Dependent Flexural Behavior of Prestressed
Reinforced Concrete Beams

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・163

By Hiroshi Watanabe, Hirohisa Koga, Hisashi Aoyama, and Yuuki Takeuchi
4-2Overnight Delivery - NJDOT Rapid Bridge Replacement
By X. Hannah Cheng and Harry A. Capers, Jr
ii

・・・・・・・・・・・・・・・・・173

4-3

Effect of Reducing Strains by SFRC Pavement on Ohira Viaduct

・・・・・・・・・・・185

By Takayoshi Kodama, Mamoru Kagata, Shigeo Higashi, Kiyoshi Itoh, andYatsuhiro Ichinose
4-4

Rapid Bridge Repair / Rehabilitation in Washington State・・・・・・・・・・・・・・・・197

By Jugesh Kapur

SESSION 5: Maintenance
5-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・209

Development of an Expansion Device for Cold Regions

・・・・・・・・・・・・・・・・・・211

By Shinya Omote, Hiroshi Mitamura, and Hiroaki Nishi
5-2

Fatigue and Corrosion in Concrete Decks with Asphalt Surfacing

・・・・・・・・・・219

By Yoshiki Tanaka, Jun Murakoshi, and Yuko Nagaya
5-3

Replacing The Suspender Ropes of a Tied Arch Bridge Using Suspension Bridge
Methods

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・233

By Barney T. Martin and Blaise A. Blabac
5-4

Suspension Bridge Cables: 200 Years of Empiricism, Analysis andManagement

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・247

By Bojidar Yanev

SESSION 6: Seismic Performance Evaluation
6-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・263

Dynamic Response Analysis of Bridge Under Seismic Loading Including
Collision・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・265

By Eiki Yamaguchi, Atsumi Ryuen, Keita Yamada, and Ryo Okamoto
6-2

Effects of Near-Fault Vertical Accelerations on Highway Bridge
Columns

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・275

By Sashi Kunnath and Huiling Zhao
6-3

Seismic Design of Multi Span Continuous RIGID-FRAME Bridge withPrestressed Concrete Box Girder・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・285

By Chiaki Nagao, Yasushi Kamihigashi, Akio Kasuga, and Kenichi Nakatsumi

iii

SESSION 7: Seismic Retrofit
7-1

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・295

Seismic Retrofit Study of CABLE-STAYED Bridge on TOKYO-GAIKANEXPRESSWAY・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・297

By Tsutomu Yoshioka, Yoshinori Kawahira, and Kouichirou Shitou
7-2

Repair of High Shear Standard Reinforced Concrete Bridge Columns
Using Cfrp

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・311

By M. "Saiid" Saiidi and Ashkan Vosooghi
7-3

Seismic Retrofit Techniques for Reinforced Concrete Bridge Columns with
Combination of FRP Sheet and Steel Jacketing

・・・・・・・・・・・・・・・・・・・・・・・・・321

By Guangfeng Zhang, Shigeki Unjoh, Jun-ichi Hoshikum, and Junichi Sakai
7-4

Development and Refinement of Illinois' Earthquake Resisting System
Strategy

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・331

By Daniel H. Tobias, James M. LaFave, Chad E. Hodel, Larry A. Fahnestock,
William M. Kramer, Joshua S. Steelman, Patrik D. Claussen, Kevin L. Riechers,
and Mark D. Shaffer

RESOLUTIONS・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・345

APPENDIX 1 : JAPAN PARTICIPANTS

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・349

Resume of Japan Participants
APPENDIX 2 : US PARTICIPANTS

・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・・357

Resume of U.S. Participantsiv

25th US-Japan
Bridge Engineering Workshop

Workshop ProgramPage Left Blank

25th U.S.-JAPAN BRIDGE ENGINEERING WORKSHOPPWRI, Tsukuba, Japan, October 19, 20 and 21, 2009
October 19, Monday − Day 1 Workshop
Conference Hall
9:00 to 9:30

Opening Session

Moderators: Masahiro Shirato and David Sanders
Welcome Remarks: Tadahiko Sakamoto (Chief Executive, PWRI)
Remarks: W. Phillip Yen (Chairman for the U.S.A Side)
Remarks: Atsushi Yoshioka (Chairman for the Japan Side)
9:30 to 9:55 Quarter Century Anniversary Special Session 1
Moderator: Atsushi Yoshioka
Speaker:
Koichi Yokoyama on behalf of the Japanese Delegation
10:20 to 12:00 Session 1: Earthquake Case Histories
Moderators: Jun-ichi Hoshikuma and W. Phillip Yen
Seismic Performance Evaluation of Seismic Retrofitted Steel Arch Bridge That Affectedby 2007 Niigata-ken Chuetsu-oki Earthquake
By Junichi Sakai and Shigeki Unjoh
Analyses of Damaged Bridge by Ground Displacement During Niigata-ken Chuetsu-oki
Earthquake
By Takao Okada and Shigeki Unjoh
Torsional Effects on Seismic Performance of Square vs. Circular RC Bridge Columns
By DJ Belarbi and S. Suriya PrakashExperimental Study on the Seismic Response of Bridge Columns Using E-Defense
By Kazuhiko Kawashima, Tomohiro Sasaki, and Koichi Kajiwara
Evaluation of the Seismic Performance of Bridge Reinforced Concrete Columns Under
Combined Actions Using Shake Table
By David Sanders and Juan G. Arias-Acosta
12:00 to 12:20 Group Photograph at the Entrance of NILIM Main Building3

13:20 to 14:50 Lab Tours in PWRI & NILIM
1. Demolished Bridge Samples
2. Research on Seismic Engineering
3. Research on Maintenance
15:10 to 16:50 Session 2: Inspection and Management
Moderators: Shoichi Nakatani and Bojidar Yanev
Analysis of Periodic Inspection Results of Highway Bridges in Japan
By Koichi Ikuta, Takashi Tamakoshi, and Masanori Okubo
Bridge Inspection and Management in California
By Barton Newton
Measures for Strategic Preventive Bridge Management of Tokyo Metropolitan Government
By Taro Awamoto and Sentaro Takagi
Instrumentation and Monitoring of I35W St. > Anthony Falls Bridge
By Catherine French, Carol Shield, Henryk Stolarski, Brock Hedegaard,
and Ben Jilk
Study for Bridge Renewal and Repair by Osaka Municipal Government
By Shinsuke Yumoto, Tetsuya Yokota, and Yasutomo Komatsu

4

October 20, Tuesday − Day 2 Workshop
Conference Hall
9:00 to 10:20 Session 3: Accelerated Bridge Construction
Moderators: Jun Murakoshi and Jugesh Kapur
Seismic Performance and Structural Details of Precast Segmental Concrete Bridge Columns
By Jun-ichi Hoshikuma, Shigeki Unjoh, and Junichi Sakai
Accelerated Constructions of the Viaducts on the Second Keihan Expressway
By Yoshihiko Taira, Hirotsugu Mizuno, Kei Muroda, and Akio Kasuga
New FHWA Seismic Hazard Mitigation Studies for Highway Bridges
By W. Phillip Yen
Overview of the Development Process and Effects of the UFO Method
By Yuji Mishima, Yukio Katsuta, and Tomoaki Tsuji
10:40 to 12:00 Session 4: Remedial Work & Partial Replacement
Moderators: Taku Hanai and "Saiid" Saiidi
Experimental Study on the Time Dependent Flexural Behavior of Prestressed
Reinforced Concrete Beams
By Hiroshi Watanabe, Hirohisa Koga, Hisashi Aoyama, and Yuuki Takeuchi
Overnight Delivery - NJDOT Rapid Bridge Replacement
By X. Hannah Cheng and Harry A. Capers, Jr
Effect of Reducing Strains by SFRC Pavement on Ohira Viaduct
By Takayoshi Kodama, Mamoru Kagata, Shigeo Higashi, Kiyoshi Itoh, andYatsuhiro Ichinose
Rapid Bridge Repair / Rehabilitation in Washington State
By Jugesh Kapur
13:20 to 14:40 Breakout Session (Conference Hall)Group A

Seismic Engineering

Moderators: Shigeki Unjoh and David Sanders
Group B

Inspection, Evaluation, and Remedial Work (CAESAR Board

Room)
Moderators: Masahiro Shirato and W. Phillip Yen

5

15:00 to 16:20 Session 5: MaintenanceModerators: Takashi Tamakoshi and Hannah Cheng
Development of an Expansion Device for Cold Regions
By Shinya Omote, Hiroshi Mitamura, and Hiroaki Nishi
Fatigue and Corrosion in Concrete Decks with Asphalt Surfacing
By Yoshiki Tanaka, Jun Murakoshi, and Yuko Nagaya
Replacing The Suspender Ropes of a Tied Arch Bridge Using Suspension Bridge Methods
By Barney T. Martin and Blaise A. Blabac
Suspension Bridge Cables: 200 Years of Empiricism, Analysis and Management
By Bojidar Yanev

6

October 21, Tuesday − Day 3 Workshop
Conference Hall
9:00 to 9:25 Quarter Century Anniversary Special Session 2
Moderator: W. Phillip Yen
Speaker:
Manos Maragakis on behalf of the US Delegation
9:25 to 10:25 Session 6: Seismic Performance Evaluation
Moderators: Junichi Sakai and Barney T. Martin
Dynamic Response Analysis of Bridge Under Seismic Loading Including collision
By Eiki Yamaguchi, Atsumi Ryuen, Keita Yamada, and Ryo Okamoto
Effects of Near-Fault Vertical Accelerations on Highway Bridge Columns
By Sashi Kunnath and Huiling Zhao
Seismic Design of Multi Span Continuous RIGID-FRAME Bridge with Prestressed
Concrete Box Girder
By Chiaki Nagao, Yasushi Kamihigashi, Akio Kasuga, and Kenichi Nakatsumi
10:45to 12:05 Session 7: Seismic Retrofit
Moderators: Yoshiki Tanaka and Catherine French
Seismic Retrofit Study of CABLE-STAYED Bridge on TOKYO-GAIKAN EXPRESSWAY
By Tsutomu Yoshioka, Yoshinori Kawahira, and Kouichirou Shitou
Repair of High Shear Standard Reinforced Concrete Bridge Columns Using Cfrp
By M. "Saiid" Saiidi and Ashkan Vosooghi
Seismic Retrofit Techniques for Reinforced Concrete Bridge Columns with Combination of
FRP Sheet and Steel Jacketing
By Guangfeng Zhang, Shigeki Unjoh, Jun-ichi Hoshikum, and Junichi Sakai
Development and Refinement of Illinois' Earthquake Resisting System Strategy
By Daniel H. Tobias, Jerome F. Hajjar, Ralph E. Anderson,
James M. LaFave, Chad E. Hodel, Larry A. Fahnestock, William M. Kramer,
Joshua S. Steelman, Patrik D. Claussen, Kevin L. Riechers, and Mark D. Shaffer712:10 to 12:30 Closing Session
Moderators: Masahiro Shirato and David Sanders
Settlement of Resolution
Remarks: W. Phillip Yen (Chairman for the U.S.A Side)
Closing Remarks: Atsushi Yoshioka (Chairman for the Japan Side)8

25th US-Japan
Bridge Engineering Workshop

Session1

Earthquake Case Histories
Seismic Performance Evaluation of Seismic Retrofitted Steel Arch Bridge That Affectedby 2007 Niigata-ken Chuetsu-oki Earthquake
By Junichi Sakai and Shigeki Unjoh
Analyses of Damaged Bridge by Ground Displacement During Niigata-ken Chuetsu-oki
Earthquake
By Takao Okada and Shigeki Unjoh
Torsional Effects on Seismic Performance of Square vs. Circular RC Bridge Columns
By DJ Belarbi and S. Suriya PrakashExperimental Study on the Seismic Response of Bridge Columns Using E-Defense
By Kazuhiko Kawashima, Tomohiro Sasaki, and Koichi Kajiwara
Evaluation of the Seismic Performance of Bridge Reinforced Concrete Columns Under
Combined Actions Using Shake Table
By David Sanders and Juan G. Arias-Acosta

Page Left Blank

SEISMIC PERFORMANCE EVALUATION OF SEISMIC RETROFITTEDSTEEL ARCH BRIDGE THAT AFFECTED BY 2007 NIIGATA-KEN
CHUETSU-OKI EARTHQUAKE
Junichi Sakai1 and Shigeki Unjoh2
Abstract
A supported deck steel arch bridge that had been retrofitted after the 1995
Hyogo-Ken Nanbu, Japan, earthquake was affected by strong ground shaking during
the 2007 Niigata-Ken Chuetsu-Oki, Japan, earthquake. The bridge suffered minor
damage by the earthquake and no structural damage of the main members wasobserved. To evaluate the seismic response during the earthquake and the effect ofseismic retrofit, a series of dynamic analyses was conducted. The results underscoredthe strengthening of the arch springing by the seismic retrofit worked to prevent theserious damage.
Introduction
The Niigata-Ken Chuetsu-Oki (Chuetsu offshore), Japan, earthquake occurredon July 16th, 2007 in the northwest Niigata region. The JMA magnitude was 6.8, andthe moment magnitude was 6.6. The VI upper of the JMA intensity was observed in the
Niigata and Nagano prefectures. The nuclear power plant was affected by theearthquake and some severe damage due to land slide was reported (NILIM, PWRI and
BRI, 2008). There was no report of severely damaged bridges by the earthquake. Somedamage of bearings and settlement of ground at bridge approaches, which were often
observed in the past earthquakes, were reported. Additionally, movement of abutment,
which was constructed on soft soil, was also reported (Okada and Unjoh, 2009).
Among the bridges that were affected by the earthquake, some bridges had
already been retrofitted and no severe damage was observed (Unjoh et al., 2008).
Although no instrumentation was provided to the bridges, it would be good examples
to evaluate the effect of seismic retrofit of bridges.
In this research, a series of dynamic analyses was conducted for a steel archbridge that had been retrofitted to evaluate the seismic response of a seismically
retrofitted bridge and the effect of the seismic retrofit.
Bridge Analyzed and Damage Due to Earthquake
Photo 1 and Figure 1 show the bridge analyzed. This is a supported deck steelarch bridge. The bridge was constructed in 1965, and the design lateral seismic
1

Senior Researcher, Bridge and Structural Technology Research Group, Center for
Advanced Engineering Structural Assessment and Research, PWRI, Japan
2
Research Coordinator for Earthquake Disaster Prevention, Research Center for
Disaster Risk Management, NILIM, Japan11

coefficient of 0.2 was used in the original design. The bridge carries two lanes of traffic
and the width is 7.5 m. The total length of the bridge, including approaches, is 197 m.
The main span is 120 m and the arch rise is 23m. Two single gerber spans are used forthe approaches.

Photo 1 Bridge analyzed
197 m
120 m

6 m 32 m

23 m

32 m 6 m

Figure 1 Bridge analyzed
Continuing Decks and Girders
at Gerber Hinge

WEST

EAST
Strengthening
Arch Crown

M

F

Unseating Prevention Device
Widening Seat Width
Strengthening Column
Providing Braces

F
Filling Concrete
into Arch Ribs

Fixing Springing

M
Unseating Prevention Device
Widening Seat Width

Strengthening Column
Providing Braces

Figure 2 Seismic retrofit

The abutments are gravity type and wall type, and the spread foundations andpile foundation are used. Two pinned bearings were used to support the main span. The
arch rib is the box type cross section with 1.5 m - height and 0.75 m- width, and the12thickness of the steel plate of the ribs for the crown region is 19mm for upper and lower
plates and 10 mm for side plates, and that for the springing region is 25 mm for upperand lower plates and 10 mm for side plates.
The seismic retrofit work was conducted in 2000 and 2001 considering theLevel 2 ground motion. Behavior after yielding of the steel members was considered in
the retrofit design. As shown in Figure 2, the items listed below were conducted as theseismic retrofit:
1. Fixing the pinned bearings at the both springings,
2. Filling the light-weight concrete into the arch rib,
3. Fixing the gerber bearings to have the bridge continuous,
4. Strengthening the arch crown,
5. Strengthening the bottom of the end columns,
6. Placing the braces for the end columns, and
7. Providing the unseating prevention devices.
The damage investigation reported that evidences of the pounding between thegirder and the abutment were observed at the deck end. It was reported that buckling
occurred at the gusset plates of lateral beams of the arch rib, but no severe damage ofthe main structural members was found (Unjoh, et al. 2008).
Analytical Model and Conditions
To simulate the seismic response during the 2007 Niigata-Ken Chuetsu-Oki
earthquake and to evaluate the effects of the seismic retrofit to the bridges, a series ofdynamic analyses was conducted for two models; one is a model for as-built bridge(As-built model), and the other is a model for the retrofitted bridge (Retrofitted model).
Although premature buckling of arch ribs was estimated to occur for the
as-built bridge, nonlinear behavior after buckling was not considered in the analyses.
Thus, the members of the arch rib were idealized as elastic beam elements. Other
structural members were also modeled as elastic elements.
Since buckling of the members of the arch ribs were not expected but theyielding of the members were expected for the Retrofitted model, nonlinear behavior
was idealized by nonlinear beam elements with the bilinear hysteretic model. The
effect of variation of axial force to the arch ribs was not considered, and the effect of
the axial force due to gravity load was only included in the analyses. Nonlinear
behavior of the members of the main girder and supporting columns were also
considered.
Table 1 compares the allowable bending moments of members at the archspringing and the arch crown. The buckling moment is shown for the allowable
moment for the As-built model, and the yielding moment is shown for the Retrofitted
model. The buckling moments at the crown were 5860 kNm and 2382 kNm for thelongitudinal and transverse directions, respectively for the As-built model, and the13allowable moment increased to 11508 kNm and 6137 kNm for the Retrofitted model
because of the infilled concrete.
Table 2 summarizes the results from the eigen-value analyses, and Figure 3
shows the mode shapes for the dominant modes for the longitudinal and transverse
directions. The deformation in the transverse direction is dominant in the 1st mode.
The natural period of the 1st mode for the As-built model is 1.6 seconds, and it
decreases to 1.16 seconds due to increment of the stiffness of the arch rib due to infilledconcrete.

Table 1 Moment capacity of arch rib (kNm)
Arch Crown
As-Built
Retrofitted

LG
5860
11508

TR
2382
6137

Springing
LG
6661
14836

TR
2229
7148

Table 2 Results from eigen-value analyses
As-built
Retrofitted
Effective Mass ratio
Natural Period Effective Mass Ratio (%) Natural Period
Mode
(sec)
(sec)
LG
TR
LG
TR
1
2
3
4
5

1.596
0.839
0.780
0.479
0.423

0
0
54.9
0
0

79.4
0
0
4.8
0

1st Mode

1.159
0.757
0.575
0.505
0.475

0
51.6
0
0
0

66.2
0
0
0
0

1st Mode

2nd Mode

3rd Mode(a) As-built model(b) Retrofitted model
Figure 3 Mode shapes

For the dynamic analyses, records obtained near the bridge during the 2007
earthquake were used, and three dimensional analyses were conduced considering two
horizontal and one vertical ground motions. The input ground motions and their
response spectra are shown in Figure 4. The design spectra of Japanese design

14

6

0

20

25

30
35
Time (sec)

40

45

2

Acceleration (m/sec )(a) Longitudinal direction
6

0

-6

15

20

25

30
35
Time (sec)

40

2

-6
15

Response Spectra (m/sec )2

Acceleration (m/sec )specifications for highway bridges (Japan Road Association, 2002) are shown here forcomparison. The peak ground accelerations are 5.95 m/sec2, 5.61 m/sec2 and 4.52
m/sec2, respectively. The horizontal records have peaks over 20 m/sec2 at around 0.25
seconds in natural period, which is larger than the design spectra, while the spectra
between 0.5 seconds and 1 second in natural periods are smaller than the design spectra.
The spectrum of the transverse direction has similar or even larger intensity than thedesign spectra around the natural period of the 1st mode while that of the longitudinal
direction has smaller intensity.

40

45

30
20

LG
TR (h = 5%)
UD
Design spectra
TypeI
TypeII100012Natural Period (sec)3(b) Transverse direction(c) Response spectra
Figure 4 Input Ground Motions

Evaluation of Effects of Seismic Retrofit
Figure 5 compares the response acceleration and displacement time histories at
the arch crown between the As-built model and the Retrofitted model. Figure 6 shows
the deformation mode when the maximum response occurred for each direction. Basedon the results for the Retrofitted model, the lateral displacements over 70 mm and 500
mm are estimated to occur in the longitudinal and transverse directions, respectively,
during the 2007 earthquake. Response acceleration increases by about 30% due tostrengthening of the bridge.
Figure 7 compares the allowable moment and the response moment. The
response moment is compared to the buckling moment for the As-built model, whilethat is compared to the yielding moment for the Retrofitted model. Based on the results
for the Retrofitted model, large bending moment occurs at the arch springings and thecrown, and the response bending moment at the crown is close to the yielding moment.

15

At the springings, the response moments are about 75% and 65% of the yielding
moment for the longitudinal and transverse directions, respectively.

As-built
Retrofitted

2

Acceleration (m/sec )20

As-built
Retrofitted

0

-20

15

20

25
30
35
Time (sec)

40

45 15

20

25
30
35
Time (sec)4045(a) Longitudinal direction(b) Transverse direction(1) Response acceleration

Displacement (mm)600

As-built
Retrofitted

As-built
Retrofitted

0

-600
15

20

25

30
35
Time (sec)

40

45 15

20

25

30
35
Time (sec)4045(a) Longitudinal direction(b) Transverse direction(2) Response displacement
Figure 5 Response of deck at arch crown

For the As-built model, response bending moment exceeds by 40% thebuckling moment at the arch crown. The response force at the bearings of the archspringings are estimated to be about 10 times larger than the horizontal and vertical
capacity of the bearings, and thus it is estimated that the bearings suffer serious
damage.
Based on the analytical results, the bearings of the arch springings could suffer
serious damage, which affects the structural stability of the arch bridge. The arch rib
could also suffer some damage at the arch crown. The responses for the Retrofitted
model do not exceed the elastic limit states, and this matches observed damage.

16

(a) Maximum response displacement in longitudinal direction (= 27.46 seconds)

15

10

5

0

-60

Yielding Moment (Retrofitted)
Response Moment (Retrofitted)

Buckling Moment (As-built)
Response Moment (As-built)
Bending Moment (MNm)

Bending Moment (MNm)(b) Maximum response displacement in transverse direction (= 26.89 seconds)
Figure 6 Deformation modes of retrofitted model

-30

0

30

15

10

5

0

-60

60

-30

0

30

60

Distance from Center (m)Distance from Center (m)(a) Longitudinal direction(b) Transverse direction
Figure 7 Maximum response moment at arch rib

Conclusions
A series of dynamic analyses was conducted for a steel arch bridge that had
been retrofitted to evaluate the seismic response of a seismically retrofitted arch bridge
and the effect of the seismic retrofit. Below are the conclusions determined from thestudy:
1. The lateral displacements over 70 mm and 500 mm are estimated to occur in17

longitudinal and transverse directions, respectively, during the 2007
earthquake.
2. The response bending moment at the crown is close to the yielding moment.
At the springings, the response moments are about 75% and 65% of theyielding moment for longitudinal and transverse directions, respectively.
The responses for the retrofitted bridge do not exceed the elastic limit states,
and this matches observed damage.
3. Without the seismic retrofit, the response bending moment exceeds the
buckling moment at the arch crown, and the response force at the bearings ofthe arch springings are estimated to be about 10 times larger than thehorizontal and vertical capacity of the bearings. The analyses underscoredthat the bearings could suffer serious damage, which affects the structural
stability of the arch bridge.
Acknowledgments
The authors would like to acknowledge the Hokuriku Regional Development
Bureau and the Nagaoka National Highway Office, MILT, for providing the
documents and materials on the bridge analyzed. The authors also appreciate the
Research Center for Disaster Risk Management, NILIM, for providing the groundmotion records.
References
Japan Road Association (2002). Specifications for highway bridges. Part V Seismic
design, Japan.
NILIM, PWRI and BRI (2008). “Report on damage to infrastructures and buildings by
the 2007 Niigata-ken Chuetsu-oki earthquake.” Technical Note of PWRI, No. 4086. (inJapanese)
Okada, T. and Unjoh, S. (2009). “Analyses of damaged bridge by ground displacement
during Niigata-ken Chuetsu-oki earthquake.” Proc. of 25th US-Japan Bridge
Engineering Workshop, Tsukuba, Japan.
Unjoh, S., Sugimoto, T., Sakai, J. and Okada, T. (2008). “Damage of bridge structuresduring recent earthquakes and the effectiveness of seismic retrofit.” Proc. of 40th Joint
Meeting US-Japan Panel on Wind and Seismic Effects UJNR, Gaithersburg, USA.

18

ANALYSES OF DAMAGED BRIDGE BY GROUND DISPLACEMENTDURING NIIGATA-KEN CHUETSU-OKI EARTHQUAKE
Takao Okada1 and Shigeki Unjoh2
Abstract
A 3-span continuous girder bridge supported by rubber bearings to disperse
seismic lateral forces was affected by 2007 Niigata-ken Chuetsu-oki earthquake. The
bridge pier columns and abutment walls were not damaged. However, all pier andabutment foundations moved with the ground displacement and some cracks were
found at the pile heads of both abutments. This paper presents the preliminary damage
analyses of the bridge to simulate the effect of the earthquake.
Introduction
The Niigata-ken Chuetsu-oki earthquake occurred on July 16, 2007. The
magnitude was 6.8 and the JMA intensity was VI upper. The depth was about 17km.
The damage was serious and a number of houses were affected. Nuclear power plant
was also affected by the earthquake. In the case of bridge structures, the similar
damages were found with those caused by the past earthquakes including damage tobearing supports, parapet walls of abutments, and settlements of backfill soils behind
the abutments. Some bridges supported by rubber bearings were affected during theearthquake. The damage of the bridges was not serious but unprecedented. The residual
displacement was found at the rubber bearings because of the ground displacement
resulting in the movement of substructures.
This paper presents the damage of the bridge and analyses to evaluate the effectof strong ground motion and ground displacement caused by this earthquake.
Overview of the Bridge and the Damages
The bridge analyzed in this study is located as shown in Fig. 1. The bridge isabout 20km away from the epicenter. The bridge was designed based on theSpecifications for highway bridges (March 2002) and was constructed in 2005 to
overpass the river as shown in Fig. 2. The bridge length is about 150m (47.6m +55m
+ 45.2m). The superstructure is 3-span continuous curved steel box girder. The
substructures are RC columns with a wall type section and supported by thecast-in-place RC pile foundations. The ground consists mainly of soft silt and clay
including thin sandy layers causing liquefaction. The ground type for seismic design isclassified in TYPE III. The pile lengths are 55 ~ 70 m and the pile diameter is 1.5m.
Bearings are rubber bearings to disperse seismic lateral forces in all direction. Two
1

Researcher, Center for Advanced Engineering Structural Assessment and Research,
PWRI
2Research Coordinator for Earthquake Disaster Prevention, Research Center for
Disaster Risk Management, NILIM

19

bearings are settled on each pier and abutment. The stoppers are installed at bothabutments in transverse direction to bridge axis. Formed cement soil structures wereused behind both abutments to decrease the earth pressure to abutments. Box culvert
structures are located behind the formed cement soil structures as shown in Fig. 3.
During the earthquake, the bridge columns and abutment walls were notdamaged, but some movements of foundations of all piers and abutments were foundas shown in Fig. 4. A2 abutment moved longitudinally by 45cm toward the river. A1
abutment moved transversely by 16cm. The superstructure rotated clockwise. Fig. 5
shows the residual displacements observed at the rubber bearings. Fig. 6 shows thejoint gap closure between the superstructure and A2 abutment. A1 abutment, P1
column and P2 column settled by 3~4 cm. The backfill soils behind abutments settled
by 25 ~ 40 cm as shown in Fig. 3. Some cracks at the pile heads of A1 and A2 abutment
were found by core drilling investigation.
After the earthquake, the bridge was repaired and retrofitted seismically. The
foundation of A2 abutment was retrofitted by adding the number of cast-in-place RCpiles. The rubber bearing were restored by jacking up the superstructure and bypushing the superstructure horizontally. The parapet wall of A2 abutment was removed
and reconstructed to restore the expansion joints and the design gap between the end of
superstructure and abutment.
Dynamic Analysis
Nonlinear dynamic analyses were performed to simulate the dynamic behavior
during the earthquake and to figure out the effect of the inertia force caused by thestrong ground motion. The analytical model used was a simple beam spring-masssystem which was commonly used in the usual seismic design. The superstructure wasmodeled by the elastic beam elements. The rubber bearings on both abutments andcolumns were modeled by the linear spring elements. The columns were modeled byelastic beam elements and nonlinear spring elements for plastic hinges. The
foundations of each substructure were modeled by the linear spring elements.
Equivalent damping ratio used in the analysis was assumed as Table 1.
The acceleration strong motion records were obtained by the seismometer ofKashiwazaki interchange which was located nearby the bridge as shown in Fig. 1.The
accelerations of three directions were used as input ground motion to perform the
dynamic analyses. The records were observed in north-south direction and east-west
direction. The input accelerations are transformed in the direction of the bridge axis
and transverse of the bridge axis. Fig. 7 shows the input accelerations and theacceleration spectra. The natural period of the bridge is 1.25 seconds. The value of theaccelerations is slightly smaller than the standard acceleration for Type II earthquake
design ground motion in the range of fundamental natural period of this bridge.
Table 2 shows the comparison of the maximum value obtained through theanalysis and the design allowable value of each structural member. The result showsthat the piers are within the elastic range and the displacement of rubber bearings are20

smaller than the design allowable displacement. The analysis estimated that the bridge
was not damaged by the inertia force by the earthquake. Since no damage was found to
the pier columns and abutment walls, the analytical result accords with the observed
situation of the bridge.
Pushover Analysis to Simulate the Bridge Movement
There is a possibility that the movement of foundations were caused by the soft
soil and liquefaction-induced ground flow. Thorough the investigation, it was clarified
the displacement of each substructure and residual displacement of rubber bearings asshown in Fig. 4 and Fig. 5. To simulate the behavior of the bridge caused by the grounddisplacement, the pushover analysis considering the observed displacement of each
substructure was performed.
The analytical model was the same with the dynamic analysis. The observed
displacement of each substructure was input to each footing of substructures tosimulate the bridge movement and the residual displacement of rubber bearings. Fig.
8 and Table 3 show the comparison of the analytical and the observed displacements.
The bridge movement and the residual displacement of the rubber bearings are
simulated well. The analytical results of directions of the A1 and P1 movement wereslightly different from the observed ones. In this analytical model, the contact of thesuperstructure with parapets of abutments was not modeled. Therefore it might affectthe difference. In fact, the superstructure contacted with both parapets and thesuperstructure was limited to move and turn around as shown in Fig. 6.
Conclusion
This paper presented the preliminaly analyses of the bridge behavior during the2007 Niigata-ken chuetu-oki earthquake. The following conclusions may be deduced:
1. The damage was not serious but the residual displacements were found at
substructures and rubber bearings which were caused by the ground displacement.
2. The dynamic analysis estimated that the bridge was not affected by the strong
ground motion. The pushover analysis considering the observed displacement ofeach substructure simulated the bridge movement and the residual displacement ofrubber bearings well.
In future study, it is needed to clarify the reason why the ground displacement
occurred. Moreover, it is necessary to consider whether rigid or elastic bearing
supports are better on the soft ground conditions including sandy layers causingliquefaction.

21

Acknowledgments
The authors would like to acknowledge the Hokuriku Regional Development
Bureau and the Nagaoka National Highway Office, MILT, for providing the
documents and materials on the bridge analyzed. The authors also appreciate East
Nippon Highway Company, for providing the ground motion records.
References
Japan Road Association (2003). Specifications for highway bridges. Part V Seismic
design, Japan.
Public Works Research Institute (2008). ‘Report on Damage to Infrastructures andBuildings by the 2007 Niigata-ken Chuetu-oki Earthquake’ Technical Note of Public
Works Research Institute, No4086, Japan
Epicenter

20km
Bridge analyzed
in this study
3.5km

Seismometer

Fig. 1 Location of the bridge analyzed

A1

P1

A2

P2

Fig. 2 Overview of the bridge supported by rubber bearings22∇ -39cm∇ -25cm
Formed cement
structures

A2

Box culvert structure

Fig. 3 Box culvert structure nearby A2 abutment
A1

P1
16cm

A2

P2
45cm

15cm
10cm

Fig. 4 Horizontal movement of A1, P1, P2 and A2
A1
17cm

A2
P1

P2

6cm

13cm

23cm
21cm

16cm

8cm

15cm

Fig. 5 Residual displacement of rubber bearings

23

Accerelation (gal)Fig. 6 Clogging of expansion joint
500

343gal

LG

0

-500
10

20

40

50

60

1800
standard
LG
TR
UD

1600

500

1400

TR

1200

0

1000

455gal-500
0Accerelation (gal)30Time (sec.)

Accelaration response
spectra (gal)Accerelation (gal)

0

10

20

30
Time (sec.)

40

50

60

800
600
400
200

500

0

UD

0
0

1

2
3
Natural period (sec.)

343gal

-500
0

10

20

30
Time (sec.)

40

50

60

Fig. 7 Input accelerations and acceleration response spectra
A1

P1

P2

A2

21cm
24cm

16cm
20cm

18cm

26cm

14cm
25cm
Fig. 8 Comparison of displacement of superstructures
Analysis :Observed data :

24

4

5

Table 1 Equivalent damping ratio for each structural element
Structural
SuperRC (except Plastic hinge
RubberFoundation
members
structure
plastic hinge) of P1 and P2
bearing
Damping
0.02
0.05
0.02
0.04
0.2
ratio
Table 2 Result of dynamic analysis
Rubber bearings
Displacement
Analysis result (m)Allowable value (m)A1longitudinal axis
0.28
0.58
transverse axis
P1longitudinal axis
0.20
0.5Transverse axis
0.15
P2
longitudinal axis
0.20
0.48
transverse axis
0.17
A2
longitudinal axis
0.30
0.51
transverse axis
Piers
Ductility of plastic hinge
Analysis result
Allowable value (µa)
P1longitudinal axis
0.81
4.3transverse axis
0.26
3.3
P2
longitudinal axis
0.84
4.2
transverse axis
0.24
3.2
Shear stress
Analysis result (kN) Allowable value (kN)P1longitudinal axis
7845
203044transverse axis
7854
27488
P2
longitudinal axis
8075
20344
transverse axis
7137
27488
Table 3 Comparison of the residual displacement of rubber bearings
Residual displacement
Analysis result (mm)Observed value (mm)of rubber bearings
G1
G2
G1
G2
A1
longitudinal axis
150
150
170
160
transverse axis00P1longitudinal axis
7
21
60
80
Transverse axis
-5
-16
-10
-20
P2
longitudinal axis
11
36
80
110
transverse axis
-64
-76
-100
-100
A2
longitudinal axis
-246
-247
-230
-210
transverse axis
10
10

25

blank26

TORSIONAL EFFECTS ON SEISMIC PERFORMANCE OF
SQUARE vs. CIRCULAR RC BRIDGE COLUMNS
A. Belarbi1, Q. Li2, and S. Suriya Prakash3
Abstract
Reinforced concrete (RC) bridge columns could be subjected to combined
flexural, axial, shear, and torsional loading during earthquake excitations. This
combination of seismic loading can result in complex flexural and shear failure ofbridge columns. Several researchers have investigated and proposed various models
for predicting seismic performance; however, knowledge of the interaction between
flexure, shear, and torsion in RC bridge columns is still limited. An experimental
study is being conducted at Missouri S&T to understand the behavior of circular andsquare RC columns under combined loading including torsion. The main variablebeing considered is the ratio of torsion-to-bending moment (T/M). The differences inbehavior between RC columns of square and circular cross section under combined
loading are discussed in this paper. The main difference between the behavior of
circular and square sections under combined loadings including torsion is related toconfinement characteristics due to transverse reinforcement arrangement as well as
warping effect in square cross sections due to torsion. In particular, the effect of crosssectional shape on hysteretic torsional and flexural response, damage distribution, andductility characteristics under combined flexure and torsional moments are discussed.
Introduction
RC bridge columns can be subjected to multi-directional ground motions
which result in the combination of axial force, shearing force, flexural and torsional
moments. The addition of significant torsion is more likely in skewed or horizontally
curved bridges, bridges with unequal spans or column heights, and bridges withoutrigger bents. In addition, structural constraints due to a rigid decking, movement ofjoints, abutment restraints, and soil conditions also lead to combined loading effects.
This combination of seismic loading can result in complex flexural and shear failureof bridge columns. Moreover, the cross-sectional details also affect the seismic
behavior of RC bridge columns, such as damage distribution and ductility
characteristics. The effect of cross section on the behavior of RC columns under
combined loading including torsion is investigated. Test results of four square andfour circular columns under cyclic flexure and shear, pure cyclic torsion, andcombined cyclic flexure and shear and torsion are presented and discussed.

1

Distinguished Professor, 2,3 PhD Candidate, Missouri University of Science &Technology, USA.

27

Previous Research
Several experimental studies have been performed to investigate the behavior
of columns under bending and shearing with and without axial compression. There
are rational and accurate models available for analyzing the interaction between axial
compression and flexure for columns. Park and Ang (1985), Priestly and Benzoni(1996), Priestly et al. (1998) and Lehman et al. (1998) have all investigated andproposed various models for predicting the seismic response under flexure. However,knowledge of the interaction between flexural and torsional moments in behavior of
RC bridge columns is limited. Few researchers have investigated the effects ofcombined loading on the seismic performance of bridge columns. The effect of
combined flexure and torsion with compression has also not been studied intensively,
and most of the tests were focused on static monotonic loads. Otsuka et al. (2004)
conducted cyclic loading tests on nine rectangular RC columns under pure torsion,
flexure and shear and different ratios of combined flexural and torsional moments.
The authors found that the hysteresis loop of torsion was significantly affected by thespacing of the transverse reinforcement. Tirasit and Kawashima (2005) tested
reinforced concrete columns under combined cyclic flexure and torsion with threedifferent rotation-to-drift ratios and formulated a nonlinear torsional hysteretic model.
The authors concluded that the flexural capacity of reinforced concrete columns
decreases as the rotation-drift ratio increases and the damage tends to occur above theflexural plastic hinge region. Recently, Belarbi et al. (2008) presented a state of theart report on behavior of RC columns under combined loadings and scope for further
research. They found that the effect of degradation of concrete strength in thepresence of shear and torsional loads and confinement of core concrete due totransverse reinforcement significantly affected the ultimate strength of concrete
sections under combined loading. They also suggested developing simplified
constitutive models to incorporate softening and confinement effects. Prakash andBelarbi (2009) reported test results of several circular columns under combined
loadings with different spiral ratios and T/M ratios. They reported that the effects ofcombined loading decrease the flexural and torsional capacity and affect the failuremodes and deformation characteristics. They also concluded that the transverse
reinforcement which might be adequate from the flexural design point of view could
be inadequate under the presence of torsional loadings.
Research SignificanceA review of previous literature revealed that there have been few studies
reporting on the behavior of RC circular and square columns under combined loading.
The effect of cross-sectional shape on the interaction between flexure and torsional
moment in the presence of axial compression has not been investigated adequately.
The seismic behavior of circular and square columns is significantly different under
combined loading due to the transverse reinforcement configurations and its effect on
confinement of concrete, variation of shear stress flow, and warping effect. For
circular columns, the spirals provide significant confinement to the core concrete
which could result in higher strength. In addition, the locking and unlocking effect of
the spiral significantly affects the behavior of circular columns due to their winding
and unwinding action during cyclic loading. The spirals when unlocked during
torsional loads cause significant spalling due to the reduced confinement effect on theconcrete core. On the other hand, the locking effect of the spirals contributes more to
the confinement of the concrete core resulting in higher strength and deformational

28

capacity. However, for the square column, the efficiency of transverse reinforcement
is somewhat lesser in confining the core concrete compared to circular columns. And
there is no effect of locking and unlocking of transverse reinforcement. Thus, thebehavior of circular and square columns needs to be clearly understood to avoid
brittle failure modes under combined loading. In addition, shear stress flow due tocombined torsional moment and shear force on the square and circular section causes
difference in damage distribution and ductility characteristics. The results from thecurrent study will provide useful contributions to establishing rational interaction
diagrams for circular and square sections under combined loading and outline
differences in behavior between the two models. The above information is essential todevelop equations for interaction surfaces and design guidelines for circular andsquare RC columns subject to combined loading including torsion.
Experimental Program
Specimen Details
Half-scale test specimens were designed to be representative of typicalexisting bridge columns. Circular specimen dimensions and reinforcement layout are
shown in Figure 1(a) and (c). Each of the circular RC column specimens had adiameter of 610 mm. and a clear concrete cover of 25 mm. Sectional details of squarecolumns are shown in Figure 1 (b) and (d), which had a width of 550 mm and clear
concrete cover of 38 mm. All these specimens were fabricated in the High Bay
Structures Laboratory at Missouri University of Science and Technology (MissouriS&T). Both the circular and square columns had the same aspect ratio (H/D=6). The
total height of the circular column was 4.5 m. with an effective height of 3.66m
measured from the top of the footing to the centerline of the applied forces; the totalheight of the square columns was 4.2 m with an effective height of 3.35 m.
Table 1 Mechanical Properties of Concrete and Steel used in Columns
CIRCULAR COLUMNSPROPERTY
Compressive
Strength(f’c, MPa)
Modulus of Rupture(fcr, MPa)
SpiralReinforcement Ratio(%)
Transverse Yield
Strength (fty, MPa)
Longitudinal Yield
Strength (fly, MPa)

SQUARE COLUMNS

H/D(6)T/M(0)H/D(3)T/M(∞)

H/D(6)T/M(0.2)

H/D(6)T/M(0.4)

H/B(6)T/M(0)H/B(6)T/M(∞)

H/B(6)T/M(0.2)

H/B(6)T/M(0.4)

33.4

28.0

41.2

41.2

36.27

34.63

40.5

40.43

3.52

3.42

3.86

3.93

3.73

3.57

3.68

3.64

0.73

1.32

1.32

1.32

1.32

1.32

1.32

1.32

450

454

457

512

Typically, the axial load due to the superstructure dead load to bridge columns varies
between 5% and 10% of the capacity of the columns. Therefore, an axial loadequivalent to 7% of the concrete capacity of the columns for both circular and squarecolumns was applied during testing. For circular columns, twelve No.8 bars (25 mm
diameter) were designed as the longitudinal reinforcement. The longitudinal29

reinforcement ratio was 2.1% for all the circular specimens. Spiral reinforcement was
designed as No.4 bars (12.5 mm diameter) with the pitch of 70 mm to obtaintransverse reinforcement ratios of 1.32%. For square columns, four No.9 bars (28mm diameter) and eight No.8 bars (25 mm diameter) were employed as thelongitudinal reinforcement providing a longitudinal reinforcement ratio of 2.13%
similar to the circular columns. To achieve better confinement of the core concrete,
rectangular and octagonal No.3 rebars were used as transverse reinforcement withspacing of 83 mm. This resulted in a transverse reinforcement ratio of 1.32% similar
to circular columns. In order to compare the seismic performance of square andcircular columns, the cross sectional dimension were chosen in a way such that boththe square and circular columns would have equal flexural and torsional capacities
with similar longitudinal and transverse reinforcement ratio. Columns under
combined loadings were tested under T/M ratios of 0.2 and 0.4. The reinforcement
details of the test specimens are given in Table 1.
PVC Pipe Assembly To
Apply Axial Load

PVC Pipe Assembly To
Apply Axial Load
910mm

910

PVC Pipe to Attach
Loading Mechanism

PVC Pipe to Attach
Loading Mechanism
560mm

560mm

560mm

560mm

25mm Diameter Longitudinal

3660mm
610mm

9mm Spirals@ 76mm O.C.

B

B

28mm Diameter Longitudinal

3352mm

9mm square and oct. ties
@ 82mm O.C.

560mm

25mm Cover

25mm Cover

610mm

610mm

38mm Cover

Cover 38mm
1524mm38mm Cover

Cover 38mm
1524mm(a) End Elevation

C

28mm Diameter LongitudinalD
9mm spiral@ 76mm O.C.

Loading Direction(b) End Elevation
25mm Diameter LongitudinalB
9mm square ties
@ 82mm O.C.

A(c) Cross Section A-A

D
H
A

G
Loading Direction

E

C
F

28mm Diameter Longitudinal

9mm Oct. ties
@ 82mm O.C.

B(c) Cross Section B-B

Figure 1 Circular and Square Column Sectional Details

Material Properties
The concrete was supplied by a local Ready Mix Plant with requested 28-day
design cylinder compressive strength of 34.5 MPa. Deformed bars were used in all
specimens. The design yield strengths of transverse and longitudinal reinforcement
are supposed to be 415 MPa. Standard tests for concrete compressive strength,
modulus of rupture, and tension tests on steel coupons were conducted. The actual
material properties of the circular and square columns on the day of the testing are
given in Table 1.

30

Test Setup and Instrumentation
The axial load was applied by a hydraulic jack on top of the columns whichtransferred the load to the column via seven un-bonded high-strength pre-stressedsteel strands. The strands all ran through a duct which was made of PVC tube in thecenter of the column and anchored to a steel plate underneath the test specimen. A
target 7% axial load ratio was applied to simulate the dead load on the column in a
bridge. Cyclic torsion, uniaxial flexure and shear, and combined flexure and shear andtorsion were generated by controlling the two horizontal servo-controlled hydraulicactuators shown diagrammatically in Figure 2. Cyclic flexure and shear load weregenerated by applying equal forces in the same direction with the two actuators. Puretorsion was created by driving equal but opposite directional forces with the two
actuators. Combined cyclic torsion and flexural moments were generated by applying
different forces/displacements with each actuator. The ratio of the imposing flexural
moment to torsional moment was controlled by maintaining the ratio of the forces ordisplacements in the two actuators. A number of instruments were used to measure
the applied loads, deformations, and internal strains. The axial load in the un-bonded
pre-stressed steel strands was measured by a load cell between the hydraulic jack andthe top of the load stub. Two load cells were installed in the horizontal hydraulicactuators to measure the applied force. Electric strain gages were attached to thesurface of the longitudinal and transverse reinforcement to measure strains in order tostudy the deformation of reinforcement under different loading conditions. The strain
gauges were mounted at different heights along the whole column in various patterns
based on different loading conditions.
Strong Wall Load Cell
Hydralic Jack
Load Stub

Two HydraulicActuators
Steel Strands(Inside Column)
Test Unit

X
Anchors

Support BlocksY
Z (out)Strong Floor(a) SIDE ELEVATION

Loading Frame
ColumnZ

X

Load StubY
Hydraulic Actuators
Strong Wall(b) PLAN

Figure 2 Test Setup

31

Loading Protocol
Experimental testing was conducted in load control mode for all columns
under flexure and shear, and combined flexure, shear and torsion loading conditions
until the first yielding of the longitudinal bars. The load was applied in load control
mode for circular columns at intervals of 25%, 50%, 75%, and 100% of the predicted
yielding of the first longitudinal bar (Fy). The horizontal displacement corresponding
to yielding of the first longitudinal bar was defined as displacement ductility one(μΔ=1). The circular column under pure torsion was loaded in load control mode atintervals of 25%, 50%, 75%, and 100% of the estimated yielding of the first spiral(Ty). The rotation corresponding to yielding of the first spiral was defined as twist
ductility one (μθ=1). For square columns, the load control mode was imposed atintervals of every 10％ of the predicted yielding of the first longitudinal bar (Fy)under flexure and shear and combined flexure, shear and torsion loading conditions.
The square column under pure torsion was loaded at 10% intervals of the predicted
yielding of the first transverse bar (Ty). More loading steps prior to yielding were
implemented in the square column testing to obtain more data for establishing theinfluence of torsion on curvature of columns under flexure. After the first yield, thetests were performed in displacement control mode until the ultimate failure of thespecimens at specific levels of ductility, meanwhile controlling the desired T/M ratios
at the same time. Three cycles of loading were performed at each ductility level. The
imposed pattern of three cycles was intended to give an indication of degradation
characteristics. The loadings were applied along the direction A-C for the circularcolumns as shown in Figure 1c. The loading along the direction A-C and C-A are
defined as positive and negative cycles, respectively. Similarly, for the squarecolumns, the loadings were applied along the direction A-D as shown in Figure 1d.
The loading along the direction D-A and A-D are defined as positive and negative
cycles, respectively
Test Results and Discussions
Columns under Flexure and Shear
Circular Column: The column tested under flexure and shear began exhibiting
flexural cracks near the bottom on side A and side C after cyclically loading thecolumn to 50% of Fy. These cracks continued to grow and new cracks appeared on
both sides of the column as higher levels of ductility were reached. The coverconcrete started spalling at a drift of about 3.2%. Spacers were attached between theactuators and the column and the displacement was applied only in the A-C direction
after ductility eight. The failure mode of the specimen began with the formation of aflexural plastic hinge at the base of the column, followed by core degradation, andfinally by the buckling of longitudinal bars on the compression side at a drift of about
12.7%. The flexural hysteresis is shown in Figure 3 up to ductility level of eight. The
flexural resistance was stable between 3% and 12.7% drift with nearly a constantflexural strength of 232.8 kN. During the last cycle of loading, a longitudinal bar
started buckling while unloading, as shown in Figure 4c. The yielding zone of thelongitudinal bars was about 610 mm from the base of the column. Longitudinal bars
on sides ‘A’ and ‘C’ both reached the yield strain at the predicted ductility level one.
The spirals remained elastic until a ductility level of six, after which they yielded.
Soon after cracking and spalling at the location of the spiral gages, the spiral gages
became non-functional.

32

Displacement Ductility
-8

-6

-4

-2

0

2

4

6

8

10

300

1000

250

-14

900
700

300
200

50

100
0

0
-100

-50

-200
-300

-100

-8

-6

-4

-2

0

2

4

6

8

10

12

14
1010

First Yielding ofLongitudinal Bar

200

Ultimate Load

808

606
150

Lateral Load (kN)400

Bending Moment (kN-m)

500

100

-10

250

600

150

-12

300

800

Ultimate Load Resistance200

Lateral Load (kN)

Displacement Ductility

1100

First Yielding of Longitudial Bar

404
100
202

50
0

0

-50

-202

-400
-500

-150

-600

-100
-404
-150

-700

-200

-800

-606
-200

-900

-250

-1000
-300

-1100
-300

-225

-150

-75

0

75

150

225

300

Bending Moment (kN-m)

-10

-808

-250
-300

-1010
-300

Displacement (mm)

-250

-200

-150

-100

-50

0

50

100

150

200

250

300

Displacement (mm)

(a)

(b)

Figure 3 Hysteresis Curves under Flexure and Shear (a) Circular and (b) Square

(a)

(b)

(c)

(1) Circular Columns

(a)

(c)

(b)

(2) Square Columns
Figure 4 Failure Modes of Columns under Flexure and Shear at (a) Longitudinal Bar
Yield, (b) Ultimate, and (c) Final failure
Square Column: The flexural hysteresis of the square column is shown in Figure 3b.
The column tested under flexure and shear started to show flexural cracking near thebottom 400 mm above the footing on side AB and side CD after cyclically loading the33column to 40% of Fy. These cracks continued to develop and new cracks appeared oneach side of the column in higher position with increasing levels of ductility.
Subsequently, the cover concrete started to spall at a drift of about 2% when thecolumn was loaded to ductility three. Application of ductility levels higher than ten
were limited by the actuator stroke capacity. After this point, the displacements
applied in the A-D direction (Negative) were limited to a smaller value of 210 mm.
During the last cycle of ductility 12, almost all the longitudinal bars buckled while
unloading as shown in Figure 4c. The yielding zone of the longitudinal bars was
about 600 mm from the base of the column. Longitudinal bars on sides ‘AB’ and‘CD’ both reached the yield strain at the predicted ductility level one. The square andoctagonal transverse reinforcement remained elastic until a ductility level of eight,
after which they yielded. The failure mode of the column began with the formation ofa flexural plastic hinge with 580 mm height from the base of the column, followed bycore concrete degradation due to crushing of concrete. The column finally failed by
the buckling and breaking of longitudinal bars and rupturing of transverse bars on thecompression side at a drift of about 8%. The progressing damage of the squarecolumn is shown in Figure 4.
Columns under Cyclic Pure Torsion
In practice, pure torsion is rarely present in structural members. It usuallyoccurs in combination with other actions often flexure and shear forces. And,
understanding the behavior of members subjected to pure torsion is necessary for
generalizing the analysis of a structural member under combined loadings. However,
only very few studies have been reported on the behavior of RC circular and squaresections under pure torsion.
Circular Column: The torsional strength of a member depends mainly on the amountof transverse and longitudinal reinforcement, the sectional dimensions, and theconcrete strength. In the post peak behavior, dowel action of longitudinal bars is also
reported to significantly affect the load resistance at higher cycles of loading (Belarbi
et al., 2008). The torsional hysteresis curve of the column tested under pure torsion isshown in Figure 5a. Under pure torsional loading, significant diagonal cracks started
developing near mid- height on the column at lower levels of ductility (Figure 5a).
The cracks lengthened when the applied torsion was increased. The progressing
damage of the specimen is shown in Figure 6a. Soon after the yielding of spirals,
spalling was observed. The angle of diagonal cracks was about 40 degrees relative tothe cross section (horizontal) of the column. The post cracking stiffness was found to
decrease proportionally with increase in the cycles of loading. The locking andunlocking effect of the spirals was observed in the negative and positive loading
cycles. During the positive cycles of twisting, the spirals were unlocked which helped
to cause significant spalling and reduced the confinement effect on the concrete core.
On the other hand, during the negative cycles of loading, the spirals underwent
locking and contributed more to the confinement of the concrete core. This effect isreflected in the unsymmetric nature of the observed hysteresis loop at higher levels of
loading (Figure 5a). At higher ductility levels, the load resistance on the negative
cycles was higher than that under positive cycles of loading due to the added
confinement generated by the locking effect of the spirals. Though, the concrete
cover spalled along the entire length of the column, significant spalling led to theformation of a torsional plastic hinge near mid-height of the column (Figure 6c). The

34

damage pattern of the column under pure torsion was significantly different from thatof column under flexure and shear.
Torsioanl Twist Ductility
-8

-6

-4

-2

0

2

4

6

Torsioanl Twist Ductility

8
-8

-6

-4

-2

0

2

4

6

8

360
360

320

Circular
320

280
240

280

Ultimate Torque240

SquareFirst Spiral Yielding
Ultimate Torque

200

160

Torsional Moment (kN-m)

Torsional Moment (kN-m)

200

First Spiral Yielding

120
80
40
0
-40
-80
-120

160
120
80
40
0
-40
-80

-160

-120

-200

-160

-240

-200

-280

-240

-320

-280
-320

-360
-20

-16

-12

-8

-4

0

4

8

12

16

20

Twist (Deg)

-20

-16

-12

-8

-4

0

4

8

12

16

20

Twist (Deg)(a) Circular Column(b) Square Column
Figure 5 Torsional Hysteresis under Pure Torsion
Square Column: The torsional moment-twist hysteresis response of the column isshown in Figure 5b. The torsional moment -twist curves are approximately linear up
to cracking torsional moment and thereafter become nonlinear with a decrease in thetorsional stiffness. The post cracking stiffness decreased proportionally with increasein the cycles of loading. The torsional resistance increased significantly due to thelongitudinal reinforcement contribution at higher load levels. Finally, the square andoctagonal ties were broken in the plastic hinge zone. Under pure torsional loading,significant shear cracks started developing near mid-height on the column at lower
levels of 60% Ty. The cracks developed in length and width when the applied
torsional moment was increased. The diagonal cracks continued to form at an
inclination of 40 to 42 degrees relative to the cross section (horizontal) of the column
as the test progressed. Concrete spalling occurred at ductility one and the spalling
region appeared along the column from bottom to top when the torsion loading
reached ductility eight. At higher levels of loading, a plastic hinge formed near midheight of the column due to significant concrete spalling. The damage pattern ofcolumn under pure torsion was significantly different from that of column under
flexure and shear, which was concentrated near the middle of the column height
instead of the typical flexural plastic hinge zone within one column cross sectional
dimension from the base of the column. Typical damage progress of the columns withsquare and circular cross section under pure torsion is shown in Figure 6.

35

(a)

(b)

(c)

(1) Circular Column
(a)

(b

(c)

(2) Square Column
Figure 6 Damage of Column under Pure Torsion at (a) Yield (b) Peak Torsional
moment and (c) Overall Failure
Columns under Cyclic Combined Flexure, Shear, and Torsion
Two circular and two square columns with transverse reinforcement ratio of
1.32% were tested under combined flexure, shear and torsion by maintaining T/Mratios of 0.2 and 0.4. The results from tests on columns under flexure and shear andpure torsion were used as the benchmarks for analyzing the behavior of specimens
under combined flexure, shear, and torsion.
Circular Columns: For the two columns tested under combined flexure and torsion,
flexural cracks first appeared near the bottom of the column. The angle of the cracks
became more inclined at increasing heights above the top of the footing withincreasing cycles of loading and depending on the amount of T/M ratio. In all thecolumns, side ‘A’ of the column exhibited less damage compared to side ‘C’. The
main reason for this is that side ‘A’ always experienced smaller displacements
compared to side ‘C’ while applying the combined loading. In general, there are three
failure modes possible under combined flexure, shear, and torsion for the concrete
member reinforced with longitudinal and transverse reinforcement: completely under
reinforced (longitudinal and transverse steel yield), partially over reinforced (onlylongitudinal steel yields or only transverse reinforcement yields), and completely
over-reinforced (concrete crushing before steel yields). The flexural and torsional
hysteresis behaviors of the column are shown in Figure 7 and Figure 8. The unsymmetric nature of the flexural envelops under combined flexure and torsion is due
to both the locking and unlocking effect and the fact that one face is subjected to36higher shearing stresses because the components of shear stresses from flexure andtorsion are additive resulting in more damage and leading to less load resistance. In allcolumns under combined flexure and torsion, failure started due to severe
combinations of shear and flexural cracks leading to progressive spalling of theconcrete cover. The columns under combined loadings finally failed due to severe
core degradation followed by buckling of longitudinal bars on side ‘C’. The typicaldamage of circular column under combined flexure and torsional moments is shownin Figure 9.
Drift Δ/L (%)
-8.2

-6.56

-4.92

-3.28

-1.64

0

Drift Δ/L (%)

1.64

3.28

4.92

6.56

-9

8.2

300
Circular
T/M=0.2
T/M=0.4

-6

-4.5

-3

-1.5

0

1.5

3

4.5

6

7.5

9

300

1100

250

-7.5

1000

SquareT/M=0.2
T/M=0.4

250
825

200

200

800

600

-50
-275
-100

100
200

50
0

0

-50

-200

-100
-400

-150

Bending Moment (kN-m)

0400

Lateral Load (kN)

275
50
0

150

Bending Moment (kN-m)

550

100

-150

-550

-600
-200

-200

First Yielding of Spiral
Reinforcement-250

First Yielding of Spiral
Reinforcement

-825

-250
-300

-1100
-300

-225

-150

-75

0

75

150

225

-800

Peak Bending Moment

Peak Bending Moment
-300

-1000
-300

300

-225

-150

-75

0

75

150

225

300

Displacement (mm)Displacement (mm)(a)(b)
Figure 7 Comparison of Flexural Hysteresis Behavior under Combined
Loading(a) Circular and (b) Square
Twist/Length (Deg/m)-4.1-3.075-2.05-1.0250

1.025

Twist/Length (Deg/m)
2.05

3.075

4.1

-4.5

360

-3.375-2.25-1.125

0

1.125

2.25

3.375

4.5

360

Circular
T/M=0.2
T/M=0.4

288

SquareT/M=0.2
T/M=0.4

288

216

216

Torsional Moment (kN-m)

Torsional Moment (kN-m)

Lateral Load (kN)

150

144

72

0

-72

-144

144

72

0

-72

-144

-216

First Yielding of Spiral
Reinforcement-216

First Yielding of Spiral
Reinforcement-288

Peak Torsional Moment-288

Peak Torsional Moment

-360

-360

-15

-12.5

-10

-7.5

-5

-2.5

0

2.5

5

7.5

10

12.5

15

-15

Twist (Deg)

-12.5

-10

-7.5

-5

-2.5

0

2.5

5

7.5

10

12.5

15

Twist (Deg)(a)(b)
Figure 8 Comparison of Torsional Hysteresis Behavior under Combined
Loading (a) Circular and (b) Square

37

(a)

(b)

(c)

(a) Circular Column at T/M=0.4
(a)

(b)

(c)

(a) Square Column at T/M=0.4
Figure 9 Comparison of Damages under Combined Loading at (a) Longitudinal
Reinforcement Yield (b) Peak Torsional moment and (c) Overall Failure
Square Columns: For the two columns tested under combined flexure and torsion,
flexural cracks first appeared near the bottom of the column at 40% of the yieldstrength, which is a smaller load level compared to the flexure and shear and puretorsion columns. The flexural and torsional hysteresis behaviors of the column are
shown in Figure 7 and Figure 8. Strength and stiffness degradation were observedwith increases in the loading cycles at each ductility level for the first two cycles, butless significant difference is seen between the curves of the second and third loading
cycles. This indicates that the deterioration of the column capacities is substantial in
the first loading cycle. It is clearly shown that due to the effect of combined loading,torsional and flexural strength reduces considerably according to the applied T/Mratio as observed in the circular columns. The flexural and torsional capacities ascompared to the pure flexure and torsion tests were indeed found to decrease due tothe effect of combined loading in this column. With increasing torsional and flexural
moments, the angle of the cracks became more inclined at increasing heights above
the top of the footing with increasing cycles of loading. The side ‘BC’ of the column
exhibited more damage compared to side ‘AD’ in the column under combined flexure
and torsion. The main reason for this is that side ‘BC’ always experienced larger
shear stress compared to side ‘AD’ while applying the combined loading. The
sequence and severity of damage of concrete and reinforcement lead to three different
possible failure modes under combined flexure, shear and torsion as above for circularcolumns. Core degradation was observed up to a higher level of 1100 mm from the38base of the column as compared to 560 mm for flexure and shear loading. This showsthat the flexure and shear plastic hinge location changes due to the effect of torsion.
However, the specific location of the plastic hinge should depend on the ratio of
applied T/M ratio. The location of the plastic hinge shifted to higher location withincreasing T/M ratios. Failure of the columns under combined loading, started due tosevere combinations of shear and flexural cracks. The column finally failed due tobuckling of longitudinal bars on side ‘AB’ and ‘CD’. Typical damage characteristics
and failure sequence of the columns under combined flexure and torsion is shown in
Figure 9.
Effect of Cross Section Shape on Flexure-Torsion Interaction
Interaction diagrams between flexural and torsional moments are shown in
Figure 10. The locking and unlocking effect of the spiral was significant in circularcolumns. It can be observed that in circular columns, the torsional strength wasreached first and then flexural strength, while the square columns reached their
torsional and flexural strength simultaneously. However, the failure sequence in all
the specimens were in the order of flexural cracking, followed by shear cracking,longitudinal bar yielding, spalling of concrete cover, spiral yielding and then final
failure by buckling of longitudinal bars after severe core degradation. It should be
noted that the T/M ratio was close to the desired loading ratio in all the specimens tillpeak torsional moment. Soon after reaching the peak torsional strength, it was notpossible to maintain the desired loading ratio as torsional strength was degradingmuch faster. Further, experimental research is in progress on behavior of squarecolumns at Missouri S&T. Additional test results at different T/M ratios and further
analysis would provide valuable information on the effect of warping and its
significance on the torsion and flexural moment interaction diagrams.
500
Cicular Columns

450

400

500

T/M ratio was not constantdue to locking effect

Square Columns

450

Torsional Moments (kN-m)

Torsional Moments (kN-m)

400
350

300

250

200

150

100

350

300

250

200

150

100
50

50
0
-1000

-800

-600

-400

-200

0

200

400

600

800

0

1000

Bending Moments (kN-m)

Locking Direction-1000

Unlocking Direction

-800

-600

-400

-200

0

200

400

600

800

1000

Bending Moments (kN-m)(a)(b)
Figure 10 Torsion and Bending Interaction at Peak Torque (a) Circular and (b) Square

Summary and Concluding Remarks
An experimental study on the effect of combined cyclic flexure and torsion on the
behavior of circular and square reinforced concrete columns was presented. Based on39

the test results presented in this paper, the following conclusions can be drawn:
1) The failure of the circular and square columns under pure torsion was caused bysignificant diagonal shear cracking leading to the formation of a torsional plastic
hinge at middle height of the column. The concrete cover spalled along the fullheight of the column.
2) The existence of torsion altered the damage patterns of reinforced concrete
columns under combined loading. Due to the presence of high shear stresses fromshear force and torsional moment, the inclined crack developed significantly
resulting in early spalling of concrete cover even before the ultimate shear was
attained.
3) The square and octagonal transverse reinforcement for square columns provided
confinement to the core concrete similar to the spiral reinforcement for circularcolumns. This ensured that the square column under flexure and shear obtained
nearly the same strength as circular columns. However, their influence on
confinement of concrete core under combined loading needs to be investigatedfurther.
4) The ultimate lateral load and displacement capacity of the columns deteriorates
with increasing levels of torsion. Similarly, the decrease of T/M ratio resulted in
the deterioration of the torsional moment and the ultimate twist capacity.
5) The locking and unlocking effect of spiral reinforcement significantly affected thefailure modes of circular columns under combined torsional and bending moments.
References
1. Belarbi, A.; Prakash, S.S.; and Silva, P. (2008), “Flexure-Shear-Torsion Interaction of RC
2.
3.

4.
5.
6.
7.
8.

Bridge Columns,” Paper No. 6, Proceedings of the Concrete Bridge Conference, St Louis,MO.
Lehman, D.E., Calderone, A.J. and Moehle, J.P. (1998), “Behavior and Design of Slender
Columns Subjected to Lateral Loading”, Proceedings of Sixth U.S. National Conferenceon Earthquake Engineering, EERI, Oakland, California. 1998.
Otsuka, H. Takeshita, E., Yabuki, W. Wang, Y., Yoshimura, T. and Tsunomoto, M.
(2004), “Study on the Seismic Performance of Reinforced Concrete Columns Subjectedto Torsional Moment, Bending Moment and Axial Force”, 13th World Conference on
Earthquake Engineering, Vancouver, Canada, Paper No. 393, 2004.
Park, Y.J. and Ang, A.H.S. (1985), “Mechanistic Seismic Damage Model for Reinforced
Concrete”, ASCE Journal of Structural Engineering, V. 111, 1985, pp. 722-739.
Prakash, S.S. and Belarbi, A. (2009), “Bending-shear-torsion interaction features of RC
circular bridge columns-An experimental study”, Special Publication-SP 265, Paper No.
SP-20, ACI Publications, Symposium in honor of Prof. Hsu (in press).
Priestly, M.J.N and Benzoni, G. (1996), “Seismic Performance of Circular Columns with
Low Longitudinal Reinforcement Ratios”, ACI Structural Journal, V. 93, No. 4, 1996, pp.
474-485.
Priestly, M.J.N, Seible, F., and Calvi, G.M. (1996), “Seismic Design and Retrofit ofBridges”, John Wiley and Sons, Inc., New York, 1996, pp. 686.
Tirasit, P., Kawashima, K., and Watanabe, G. (2005), “An Experimental Study on the
Performance of RC columns Subjected to Cyclic Flexural Torsional Loading”, Second
International Conference on Urban Earthquake Engineering, Tokyo, Japan, pp. 357-364.

40

EXPERIMENTAL STUDY ON THE SEISMIC RESPONSE OF BRIDGECOLUMNS USING E-DEFENSE
Kazuhiko Kawashima1, Tomohiro Sasaki2, and Koichi Kajiwara3
Abstract
This paper presents preliminary results of a large scale shake table experiment forstudying the failure mechanism of three reinforced concrete bridge columns; a typicalflexural dominant column in the 1970s (C1-1), a typical shear failure dominant column inthe 1970s (C1-2) and a typical column designed in accordance with the current design code
(C1-5). They were 7.5 m tall 1.8-2.0m diameter circular reinforced concrete columns. They
were subjected to a near-filed grounds motion recorded during the 1995 Kobe, Japan
earthquake. Preliminary results on the experiment and analytical correlation are presented.
Introduction
Bridges are a vital component of transportation facilities; however it is known thatbridges are vulnerable to the seismic effect. Bridges suffered extensive damage in pastearthquakes such as 1989 Loma Prieta earthquake, 1994 Northridge earthquake, 1995
Kobe earthquake, 1999 Chi Chi earthquake, 1999 Bolu earthquake and 2008 Wenchuan
earthquake. A large scale bridge experimental program was initiated in 2005 in the
National Research Institute for Earth Science and Disaster Prevention (NIED), Japan as
one of the three US-Japan cooperative research programs based on NEES and E-Defense
collaboration. In the bridge program, it was originally proposed to conduct experiments on
two model types; 1) component models and 2) system models. They are called hereinafteras C1 experiment and C2 experiment, respectively (Nakashima 2008).
The objective of the C1 experiment is to clarify the failure mechanism of reinforced
concrete columns using models with as large section as possible. On the other hand, C2
experiment was proposed to clarify the system failure mechanism of a bridge consisting ofdecks, columns, abutments, bearings, expansion joints and unseating prevention devices.
C1 experiment was conducted for two typical reinforced concrete columns which
failed during the 1995 Kobe earthquake (C1-1 and C1-2 experiments) and a typicalreinforced concrete column designed in accordance with the current design requirements(C1-5 experiment). This paper shows preliminary results of the experiment and analysis on
three C1 columns.

1

Professor, Department of Civil Engineering, Tokyo Institute of Technology
Graduate Student, Department of Civil Engineering, Tokyo Institute of Technology
3
Senior Researcher, National Research Institute for Earth Science and Disaster Prevention
2

41

1500

D29@32bars
D29@32bars
D29@16bars
6000
9300

28@300=8400

D13@300mm
100
100 100
1800
1800

15@150=2250 13@300=3900 15@150=2250

Photo 1 C1 on E-Defense

7000

1500

1500

100

7000

6000
9300

D35@36bars
D35@36bars

D22@150mm
150 100
1800
1800

1800

D29@32bars
D29@32bars
D29@16bars

27@300=8100

100 100
1800

56@150=8400

6000
9300

1800

10@300=3000

15@150=2250 13@300=3900 15@150=2250
16@300=4800(a) C1-1

D13@300mm
100
100 100
1800

7000(b) C1-2(c) C1-5
Fig. 1 C1 column models
Column Models

Photo 1 shows the experimental setup of three columns using E-Defense(Kawashima et al. 2009). Two simply supported decks were set on the column and on thetwo steel end supports. A catch frame was set under the lateral beam of the column toprevent collapse of the column when it was excessively damaged. Tributary mass to the42column by two decks including four weights was 307 t and 215 t in the longitudinal andtransverse directions, respectively. The tributary mass was increased by 21 % from 307 t to
372 t in a part of C1-5 excitation.
Three full-size reinforced concrete columns as shown in Fig. 1 were constructed forthe experiment. Columns used for C1-1, C1-2 and C1-5 experiments, which are calledhereinafter as C1-1, C1-2 and C1-5, respectively, are 7.5 m tall reinforced concrete
columns with a diameter of 1.8 m in C1-1 and C1-2 and 2 m in C1-5. C1-1 and C1-2 are
typical columns which were built in the 1970s based on a combination of the static lateralforce method and the working stress design in accordance with the 1964 Design
Specifications of Steel Road Bridges, Japan Road Association. Since it was a common
practice prior to 1980 to terminate longitudinal bars at mid-heights, the inner and center
longitudinal bars were cut off at 1.86 m and 3.86 m from the column base, respectively.
The cut-off heights were determined by extending a length equivalent to a lap splicing
length l ls (about 30 times bar diameter) from the height where longitudinal bars became
unnecessary based on the moment distribution. On the other hand, longitudinal bars were
not cut-off in C1-1. C1-1 and C1-2 had the same shape, heights, bar arrangement andproperties except the cut-off. As a consequence, C1-1 failed in flexure while C1-2 failed inshear, as will be described later. The shear failure due to cut-off was one of the major
sources of the extensive damage of bridges in the 1995 Kobe earthquake (Kawashima andUnjoh 1997). Combination of the lateral seismic coefficient of 0.23 and the verticalseismic coefficient of +/-0.11 (upward and downward seismic force) was assumed in thedesign of C1-1 and C1-2.
Deformed 13 mm diameter circular ties were provided at 300 mm interval, except
the outer ties at the top 1.15m zone and the base 0.95 m zone where they were provided at
150mm interval in C1-1. Ties were only lap spliced with 30 times the bar diameter. Lap
splice was a common practice by the mid 1980s. The longitudinal and tie bars had anominal strength of 345 MPa (SD345), and the design concrete strength was 27 MPa. The
longitudinal reinforcement ratio Pl was 2.02 % and the volumetric tie reinforcement ratio
ρ s was 0.32 % except the top 1.15 m and base 0.95m zones where ρ s was 0.42% in C1-1.
Pl and ρ s varied depending on the zones in C1-2; 2.02 % and 0.42 % at the base 0.95 mzone, 2.02 % and 0.32 % between 0.95 m and 1.86 m, 1.62 % and 0.21 % between 1.86 m
and 3.86 m, 0.81 % and 0.11 % between 3.86 m and 4.85 m, and 0.81 % and 0.21 % at the
top 1.15 m zone, respectively.
On the other hand C1-5 was designed in accordance with the 2002 JRA Design
Specifications of Highway Bridges (JRA 2002). Sixty four deformed 35mm diameterlongitudinal bars were provided in two layers. Deformed 22 mm diameter circular ties
were set at 150 mm and 300 mm interval in the outer and inner longitudinal bars,
respectively. The ties were developed in the core concrete using 135 degree bent hooks
after lap spliced with 40 times the bar diameter. The nominal strength of longitudinal andtie bars and the design concrete strength were the same with those in C1-1 and C1-2

43

columns. The longitudinal reinforcement ratio Pl was 2.19 % and the volumetric tie
reinforcement ratio ρ s was 0.92 %Evaluation of the seismic performance of C1-1 and C1-5 in the longitudinal
direction based on the 2002 JRA code is as follows: Because the design response
acceleration S A is 17.16 m/s2 for both C1-1 and C1-5, the yield displacement u y andultimate displacement u u are 0.046 m and 0.099 m in C1-1 and 0.045 m and 0.231 m in
C1-5. The design displacement u d is evaluated from u y and u u as

ud = u y +uu − u y(1)αin which α depends on the type of ground motion (near-field or middle field groundmotion) and the importance of the bridge. Assuming α is 1.5 for a combination of thenear-field ground motion category and the important bridges category, the design
displacement u d is 0.081 m in C1-1 and 0.169 m in C1-5.
On the other hand, the displacement demand u is 0.328 m in C1-1 and 0.168 m inC1-5 because the force reduction factor is 1.58 and 2.56 respectively. Consequently, C1-1
and C1-5 were evaluated to be unsafe and safe, respectively based on the current designcode.
Acceleration
2(m/sec )

10
5
0
-5

Table Motion
Target-10(a) Longitudinal
Acceleration
2(m/sec )

10
5
0
-5

Table Motion
Target-10
Acceleration
2(m/sec )(b) Transverse
5

Table Motion

0
-5
0

5

10
Time (sec)Target1520(c) VerticalFig. 2 100% E-Takatori ground motion (C1-5(1)-1 excitation)

44

Three columns were excited using a near-field ground motion as shown in Fig. 2
which was recorded at the JR Takatori Station during the 1995 Kobe earthquake. It was
one of the most influential ground motions to structures. However duration was short.
Taking account of the soil structure interaction, a ground motion with 80% the original
intensity of JR Takatori record was imposed as a command to the table in the experiment.
This ground motion is called hereinafter as the 100 % E-Takatori ground motion.
Excitation was repeated to clarify the seismic performance of the columns when they were
subjected to near-field ground motions with longer duration and/or stronger intensity. Only
C1-5 was excited using 125 % E-Takatori ground motion with 21 % increased deck mass
to study the seismic performance under a stronger ground motion than the JR-Takatori
Station ground motion.

Seismic Performance of C1-1 and C1-5
Progress of failure
C1-1 was subjected to the 100 % E-Takatori ground motion twice. Photo 2 shows
the progress of failure at the plastic hinge on the SW surface where damage was mostextensive. NS and EW direction correspond to the transverse and longitudinal directions,
respectively, of the model. During the first excitation (C1-1-1 excitation), at least two outerlongitudinal bars from S to W locally buckled between the ties at 200 mm and 500 mm
from the base. During the second excitation (C1-1-2 excitation), both the covering and coreconcrete suffered extensive damage between the base and 0.7 m from the base on the SW
surface. Three ties from the base completely separated at the lap splices. Eleven outer andthree center longitudinal bars locally buckled between ties at 50 mm and 500 mm from thebase.

(a) C1-1-1 excitation (8.35s)(b) C1-1-2 excitation (7.71s)

Photo 2 Progress of damage of C1-1
On the other hand, C1-5 was subjected to the 100% E-Takatori ground motion
twice (C1-5(1)-1 and C1-5(1)-2 excitations). After the mass was increased by 21 % from307 t to 372 t, C1-5 was subjected to the 100% E-Takatori ground motion once (C1-5(2)45excitation). Then C1-5 was subjected to the 125% E-Takatori ground motion twice(C1-5(3)-1 and C1-5(3)-2 excitations).
Photo 3 shows the progress of failure of C1-5 at the plastic hinge during C1-5(1)-1,C1-5(2) and C1-5(3)-2 excitations. During C1-5(1)-1 excitation, only a few flexural cracks
with the maximum width of 1mm occurred around the column at the plastic hinge.
Therefore it is noted that the seismic performance is enhanced in C1-5 than C1-1 under the
first 100% E-Takatori excitation. The damage progressed during C1-5(2) excitation suchthat the covering concrete spalled off at the 500 mm base zone from WSW to SSW. During
C1-5(3)-2 excitation, the failure extensively progressed. The core concrete crashed due torepeated compression, and blocks of crashed core concrete spilled out from the steel cages
like explosion. Such a failure was never seen in the past quasi-static cyclic or hybrid
loading experiments. Because the maximum aggregate size was 20 mm, the concrete
blocks after crashed can be as small as 20-40 mm. Because the gaps of longitudinal bars
and circular ties were 132 mm and 128 mm, respectively, it was possible for the blocks ofcrashed core concrete to move out from the steel cages. Furthermore twelve outerlongitudinal bars and nineteen inner longitudinal bars locally buckled on SW and NE-E
surfaces. The 135 degree bent hooks developed in the core concrete still existed in theoriginal position although the core concrete around the hooks suffered extensive damage.

(a) C1-5(2) excitation (8.80s)(b) C1-5(3)-2 excitation (7.17s)

Photo 3 Progress of damage of C1-5
Response displacement and moment capacity
Figs. 3 and 4 show the response displacement at the top of C1-1 and C1-5,
respectively, in the principal response direction (nearly SW-NE direction). The peak
displacement of C1-1 was 0.179 m (2.4 % drift) during C1-1-1 excitation while the peak
displacement of C1-5 was 0.084 m (1.1 % drift) during C1-5(1)-1 excitation. Because theultimate displacement in accordance with JRA 2002 code was 0.100 m and 0.235 m in
C1-1 and C1-5, respectively, the above peak response displacements corresponded to
179 % and 36 % the ultimate displacement in C1-1 and C1-5, respectively.

46

Figs. 5 and 6 show the moment at the column base vs. lateral displacement at thecolumn top hystereses of C1-1 and C1-5, respectively, in the principal response direction.
The computed moment vs. lateral displacement relations based on the 2002 JRA code arealso shown here for comparison. The moment capacity of C1-1 during C1-1-2 excitation
was 13.41 MNm which deteriorated by 19 % from the moment capacity during C1-1-1
excitation of 16.47 MNm. On the other hand, the moment capacity of C1-5 column
progressed from 19.82 MNm during C1-5(1)-1 excitation to 20.14 MNm and 24.85 MNmduring the C1-5(2) and C1-5(3)-2 excitations, respectively. However since the moment
capacity of C1-5 during the C1-5(3)-1 excitation was 25.54 MNm, the moment capacity of
C1-5 deteriorated by 3% during C1-5(3)-2 excitation. The computed moment capacities
are close to the experimental values in both C1-1 and C1-5, however the computed
ultimate displacement are very conservative compared to the experiment.

0.3

Displacement (m)Displacement (m)0.6

C1-1-1C1-1-2

0.2
0.1
0
-0.1

0.4
0.3
0.2
0.1
0

-0.2
0

5

10
Time (sec)

15

-0.1
0

20

20
u
10

u

y

uu

d u

y0

0
-10
-20

C1-1-1C1-1-2
JRA 2002-0.2 -0.1 0 0.1 0.2 0.3
Lateral Displacement (m)510Time (sec)15

20

Fig. 4 Response displacement at the topof C1-5 in the principle response
direction

Moment at the base (MNm)

Fig. 3 Response displacement at the topof C1-1 in the principle response
direction
Moment at the base (MNm)

C1-5(1)-1C1-5(2)C1-5(3)-2

0.5

u

20
u

u u
d

u

y

y0

10
0
C1-5(1)-1C1-5(2)C1-5(3)-2
JRA 2002-20-0.3 -0.2 -0.1 0 0.1 0.2 0.3 0.4 0.5 0.6 0.7
Lateral Displacement (m)-10Fig. 5 Moment at the base vs. lateral Fig. 6 Moment at the base vs. lateral displacement
displacement of the topof the top hysteresis of C1-5 in the principle
hysteresis of C1-1 in the
response direction
principle response direction

47

(a) NW(b) SE(1) 6.50s(a) NW(b) SE(2) 6.87s

Photo 4 Progress of damage of C1-2
Seismic Performance of C1-2
Progress of frailure
Photo 4 shows the progress of failure of C1-2 on NW and SE surfaces. A horizontal
crack first developed at 4.10s along NW to E surface, and it progressed to a shear crack at
4.33 s. Another horizontal crack developed at 4.60s along W to SE surface, and it extended
to at least two diagonal cracks at 4.87s. Among two diagonal cracks developed at 4.33 s,a crack on NW surface extended to W surface, and the other crack on SE surface extended
to S at 5.37s. The core concrete started to crash due to shear, and the blocks of crashed coreconcrete started to move out from the inside of the column near the upper cut-off on N and
NW surfaces at 6.04 s. The same but more extensive failure occurred on S and SW surfaces
at 6.504 s. The blocks of crashed core concrete progressively moved out from steel cages
associated with the column response in the SW direction.
At 6.87 s, the bottom of lateral beam hit with the upper surface of catch frame due48

to excessive response displacement. Three circular tie bars completely separated at theirlap splice and the longitudinal bars deformed in the outward direction. Extensive failure ofcore concrete and deformation of longitudinal bars progressed on W, NW, N, NE and E
surfaces.
It should be noted in the above process that the failure of core concrete was
extensive and a large numbers of blocks of crashed core concrete as well as deformed
longitudinal bars moved out from inside of the column during very short time (less than 3s). It was like an explosion.
Response and shear capacity

Displacement (mm)Fig. 7 shows response displacement of C1-2 in the principal response direction. As
described above, since bottom of the lateral beam hit with the upper surface of catch frame
at 6.87 s, the column response after 6.87 s was affected by this contact. Without the catch
frame, the column possibly overturned. Therefore the response displacement after thiscontact is plotted by dotted line in Fig. 7. At 7.125 s, right after the contact, the column
response displacement reached its peak of 439.2 mm and 253.0 mm in the longitudinal andtransverse directions, respectively. Residual drifts of 204.5 mm and 343.2 mm weredeveloped after the excitation.
0.5
0.4
0.3
0.2
0.1
0
-0.1
-0.2
-0.3
0

f

a

c
e
b d

5

10
Time (sec)15

20

Fig. 7 Response displacement at the top C1-2 in the principle response direction
Fig. 8 shows the lateral force at the upper cut-off vs. lateral displacement at thecolumn top hysteresis in the principal response direction. The hysteresis after the contactof the column with the catch frame is plotted by dotted line. The shear capacity of thecolumn Fs was evaluated based on the truss theory (Priestly 1996).
The shear stress at the upper cut-off vs. the lateral displacement at the column top
relation was evaluated as shown in Fig. 9, in which τ c is normalized in terms of α c andα pl defined as49

⎛ 24 ⎞
α c = ⎜⎜ ⎟⎟⎝ fc ⎠−1 / 3⎛ 0.012 ⎞⎟⎟; α pl = ⎜⎜⎝ pl ⎠−1 / 3

(2)

3000

1.5

2000

1
τ / α α (MPa)

0.5

c pl

1000
0
-1000

0

c

Lateral Force (kN)In Fig. 9, shear stress evaluated for two 1.68m tall 400mm diameter scaled modelcolumns with different shear vs. flexure strength ratio is included for comparison (Sasaki
et al. 2008). τ c / α c α pl of C1-2 is 0.68 MPa which is 15% larger than the value evaluatedbased on the shear equation (0.59 MPa) by Kono et al (Kono et al. 1996).

-0.5

-2000
-1

-3000

C1-2
Scale model (γ=0.84)
Scale model (γ=0.74)-1 -0.5 0 0.5 1 1.5 2 2.5 3
Lateral Displacement (m)

-0.2
0
0.2
0.4
Lateral Displacement (m)Fig. 8 Lateral force at upper cut-off vs. lateraldisplacement at the column top hysteresis
in the principle response direction

Fig. 9 Shear stress of concrete

Analytical Correlation for C1-5

Analytical correlation for C1-5 during C1-5(2) and C1-5(3)-2 excitations is shown
here. The column was idealized by a 3D discrete analytical model including P - ∆ effectas shown in Fig. 10. The column was idealized by fiber elements. A section was divided
into 400 fibers.

Fig. 10 Analytical model
The stress vs. strain constitutive model of confined concrete is assumed based on

50

Hoshikuma et al (1997) and Sakai and Kawashima (2006). The Modified Menegotto-Pinto
model was used to idealize the stress vs. strain relation of longitudinal bars (Menegotto andPinto 1973, Sakai and Kawashima 2003).

2

0.1
0.05
0
-0.05
-0.1
0

Acceleration (m/sec )Displacement (m)Fig. 11 shows the analytical correlation on the response displacements at the top of
the column in the principal direction during C1-5(1)-1 and C1-5(2) excitations. Fig. 12
compares the measured and computed moment at the base vs. lateral displacement at thecolumn top hystereses during the two excitations. Because nonlinear hysteretic response
was still limited during C1-5(1)-1 excitation, the computed response displacement andmoment vs. lateral displacement hysteresis are quite in good agreement with theexperimental results, however as C1-5 suffered more damage, the accuracy of analyticalprediction decreases.

5

10Time (sec)Analysis
Experiment
15
20

15
10
5
0
-5
-10
-15
0

(a) Displacement

5

10Time (sec)Analysis
Experiment
1520(b) Acceleration2

0.3
Analysis
Experiment

0.2
0.1
0
-0.1
0

Acceleration (m/sec )Displacement (m)(1) C1-5(1)-1 excitation

5

10
Time (sec)

15

20

15
10
5
0
-5
-10
-15
0

(a) Displacement

5

10Time (sec)Analysis
Experiment
1520(b) Acceleration(2) C1-5(1)-2 excitation

Fig. 11 Analytical correlation for the response displacement and acceleration at thecolumn top in the principle response direction

Consequently, it is required to develop an analytical model that can predict the
response of the columns until collapse for realizing reliable performance based seismic
design.

51

y

u u
d

Moment at the base (MNm)

Moment at the base (MNm)

u

20

u

u
10

y0

0
-10
Analysis
Experiment-20
JRA 2002-0.3 -0.2 -0.1 0 0.1 0.2 0.3
Lateral Displacement (m)(a) C1-5(1)-1 excitation

u

20

y

u

u

d

u

u

y0

10
0
-10
Analysis
Experiment
JRA 2002-0.3 -0.2 -0.1 0 0.1 0.2 0.3
Lateral Displacement (m)-20(b) C1-5(2) excitation

Fig. 12 Moment vs. lateral displacement at the column top in the principle response
direction
Conclusions

A preliminary result on a series of shake table experiment and analysis to threefull-size reinforced concrete columns was presented. Based on the results presented herein,
the following tentative conclusions may be deduced;
1) C1-1 which is a typical column in the 1970s suffered extensive damage underC1-1-1 excitation. The progress of damage during C1-1-2 excitation was extensive eventhough it was anticipated before the experiment that damage would not progress unless theintensity of second excitation was much larger than that of the first excitation. This resulted
from the extensive deterioration of the lateral confinement due to separation of ties at thelap splices. It is highly possible that columns without sufficient lateral confinement havea similar progress of damage during a long-duration near-field ground motion or strong
aftershocks.
2) C1-5 which is a typical column in accordance with the current design criteria
suffered only a few numbers of horizontal cracks with the maximum width of 1 mm underC1-5 (1)-1 excitation. The ultimate drift was 2.9 % which was 2.2 times larger than that of
C1-1. Consequently, enhancement of the seismic performance of C1-5 compared to C1-1
is obvious. However the progress of failure of C1-5 was extensive when it was subjected to
25 % stronger excitation under 21% added mass (C1-5(3) excitations). Blocks of crashedcore concrete spilled out like explosion from the steel cages. The seismic performance ofC1-5 subjected to longer duration near-field ground motion has to be carefully evaluated.
3) C1-2 failed in shear at the upper cut-off. As soon as circular ties at the upper
cut-off yielded, a small diagonal cracks developed. As they extended to several major
diagonal cracks, C1-2 completely failed in shear within less than 2.5 s since the initiationof a couple of small diagonal cracks. Concrete blocks crashed by shear and deformed52

longitudinal bars extensively moved out from the inside of column.
4) The lateral confinement in the flexure dominant columns is not uniform aroundthe ties as it is currently assumed in design. More importantly, the lateral confinement ofmulti layered ties is very complex. Strains of ties are not similar among the multi-layered
ties, and they are related to the degree of constraint exerted for preventing local bucklingof longitudinal bars. Strains are generally larger in the outer ties than the inner ties. Thisimplies that the lateral confinement by Eq. (2) can be overestimated.
5) Computed response for the flexure dominant columns is satisfactory while
response undergoes the moderate nonlinear range, however accuracy of the analyticalprediction deteriorates once the columns undergo the strong nonlinear range. An analyticalmodel which can predict response of the columns until failure should be developed forenhancing the reliability of the performance based seismic design.

Acknowledgments

The C1 experiment was conducted based on the extensive support of over 70
personnel in the Overview Committee (Chair, Professor Emeritus Hirokazu Iemura, Kyoto
University), Executing Committee of Large-scale Bridge Experimental Program (Chair,
Professor Kazuhiko Kawashima, Tokyo Institute of Technology), Analytical Correlation
WG (Chair, Dr. Shigeki Unjoh, Public Works Research Institute), Measurements WG(Chair, Professor Yoshikazu Takahashi, Kyoto University), Dampers & Bearings WG(Chair, Dr. Masaaki Yabe, Chodai) and Blind Analysis WG (Chair, Professor Hiroshi
Mutsuyoshi, Saitama University). Their strong support is greatly appreciated. Invaluable
support and encouragement of Professor Stephen Mahin, University of California,
Berkeley and Professor Ian Buckle, University of Nevada, Reno are greatly appreciated.

References

Nakashima, M., Kawashima, K., Ukon, H. and Kajiwara, K. (2008), “Shake table
experimental project on the seismic performance of bridges using E-Defense,”
S17-02-010 (CD-ROM), 14 WCEE, Beijing, China.
Hoshikuma, J., Kawashima, K., Nagaya, K. and Taylor, A.W. (1997), “Stress-strain model
for confined reinforced concrete in bridge piers,” Journal of Structural Engineering,
ASCE, 123(5), 624-633.
Kawashima, K. Sasaki, T., Kajiwara, K., Ukon, H., Unjoh, S., Sakai, J., Takahashi, Y.,
Kosa, K., and Yabe, M. (2009), “Seismic performance of a flexural failure typereinforced concrete bridge column based on E-Defense excitation,” Proc. JSCE, A,
Invited paper, 1-19, JSCE.
Kawashima, K. and Unjoh, S. (1997), “The damage of highway bridges in the 1995

53

Hyogo-ken nanbu earthquake and its impact on Japanese seismic design,” Journal of
Earthquake Engineering, 1(3), 505-542.
Kono, H., Watanabe, H. and Kikumori, Y. (1996), “Shear strength of large RC beams,”
Technical Note, 3426, Public Works Research Institute, Tsukuba, Japan.
Priestley, M.N.J., Seible, F. and Calvi, G.M. (1996), “Seismic design and retrofit ofbridges,” John Wiley & Sons, New York, USA.
Menegotto, M. and Pinto, P.E. (1973), “Method of analysis for cyclically loaded R.C.
plane frames including changes in geometry and non-elastic behavior of elements under
combined normalized force and bending,” Proc. IABSE Symposium on Resistance andUltimate Deformability of Structures Acted on by Well Defined Repeated Loadings,
15-22.
Sakai, J. and Kawashima, K. (2003), “Modification of the Giuffre, Menegotto and Pinto
model for unloading and reloading paths with small strain variations,” Journal of
Structural Mechanics and Earthquake Engineering, JSCE, 738/I-64, 159-169.
Sakai, J. and Kawashima, K. (2006), “Unloading and reloading stress-strain model for
confined concrete,” Journal of Structural Engineering, ASCE, 132 (1), 112-122.
Sasaki, T., Kurita, H., and Kawashima,K. (2008), “Seismic performance of RC bridge
columns with termination of main reinforcement with inadaquate development,”
S17-02-009 (CD-ROM), 14WCEE, Beijing, China.

54

EVALUATION OF THE SEISMIC PERFORMANCE OF BRIDGE REINFORCED
CONCRETE COLUMNS UNDER COMBINED ACTIONS USING SHAKE TABLE
Juan G. Arias-Acosta1 and David H. Sanders2
Abstract
Combined actions (axial, shear, bending and torsion) can have significant effects on
the force and deformation capacity of reinforced concrete bridge columns (RCC); these load
combinations can result in unexpected large deformations and extensive damage that canseriously affect the seismic performance of bridges. To study the impact of different loadings
combinations on both circular and non-circular sections (double interlocking spirals), eightlarge-scale cantilever-type RCC specimens will be tested on the bidirectional shake tablefacility at University of Nevada, Reno (UNR). As part of the study, an inertial loading system
named Bidirectional Mass Rig was developed to allow shake table testing of single RCC
under biaxial ground motions. Two sets of circular and interlocking RCC will be subjected todifferent levels of biaxial, torsion and vertical loads through real time earthquake motions.
The performance of the specimens will be assessed in terms of strength, deformation, andfailure mode.
Introduction
During moderate to large earthquakes, reinforced concrete bridge columns are
subjected to combinations of actions and deformations, caused by spatially-complex
earthquake ground motions, structural configurations and the interaction between input andresponse characteristics. As a result, the seismic behavior of RCC will be seriously affected,
and that in turn influences the performance of bridges as essential components oftransportation systems. In addition, current analysis methods, behavior theories and design
practices do not take into consideration the full range of interactions, due to the scarcity ofexperimental data and a lack of behavioral understanding.
In order to address the complex behavior of bridge members under combined loadings
and its impact on system response, a comprehensive project sponsored by the National
Science Foundation was established in 2006. This project includes researchers from six
institutions, and the objectives are to develop a fundamental knowledge of the impact of
combined actions on column performance and their implications on system response throughanalytical and experimental research.
The work at UNR focuses on the development of refined analysis and shaking table
tests of cantilever-type scale models of bridge columns subjected to different levels of
biaxial, torsion and vertical loads through real time earthquake motions. The performance ofthe specimens will be assessed in terms of strength, deformation, energy dissipation andfailure mode. These results will be used to validate analytical tools, developing new inelastic
models for RCC under combined loadings and to propose new design methodologies. This
paper highlights some of the preliminary work underway at UNR.

1
2

Graduate Student, Dept. of Civil & Environmental Engineering, Univ. of Nevada, Reno, NV89557, USA
Professor, Dept. of Civil & Environmental Engineering, Univ. of Nevada, Reno, NV89557, USA

55

Specimen Details
Two sets of specimens of circular and double interlocking columns were constructedusing current bridge design details typical of bridges in California in accordance with the
Seismic Design Criteria (CALTRANS, 2006). The structural configuration selected was
similar to previous columns tested at UNR (Laplace et al., 1999 and Correal et al., 2004). For
circular columns the scaling factor selected was 1/3. The diameter of the specimens was 406
mm (16 in) and the height 1830 mm (72 in), thus the aspect ratio was 4.5, which allows for
flexural dominated behavior. The columns were reinforced with 20 No.4 (D13) deformed
longitudinal bars, distributed uniformly around the perimeter and fully developed with 90
degree hooks in the footing. This resulted in a longitudinal reinforcement ratio of 2%. The
confinement consisted of a continuous spiral made from galvanized steel wire with adiameter of 6.25 mm (0.25 in) and a pitch of 38 mm (1.5 in). The clear cover was set to 19
mm (0.75 in) and the resulting volumetric ratio of the spiral reinforcement was 0.92%.
Details of circular specimens are shown in Fig. 1.
For double interlocking columns a scale factor of 1/4 was used. Consequently, the heightwas 1830 mm (72 in) and the width in the short side was 305 mm (12 in), while that in thelong dimension was 445 mm (17.5 in). The longitudinal reinforcement consisted of 32 No. 3(D10) deformed bars, spaced evenly in two circular patterns and fully developed in thefooting. The resulting reinforcement ratio was 2%, while the volumetric ratio of the spiralreinforcement was 1.0%. The confinement for each of the circular sections consisted of acontinuous spiral made from galvanized steel wire with a diameter of 4.9 mm (0.192 in) and apitch of 25 mm (1.0 in). The clear cover was set to 13 mm (0.5 in). Details of the interlocking
specimens are shown in Fig. 2.
The design compressive strength of the concrete was set as 30 MPa (4.5 ksi), while thenominal yielding strength of the steel was 447 MPa (64 ksi) for deformed bars and 420 MPa(60 ksi) for steel wire. Table 1 shows the real properties of steel and concrete based oncoupons and cylinders tests. The superstructure mass was defined as 356 kN (80 kips), whichis equivalent to an axial load of 8% of Agf’c.
Experimental Test Setup
As part of the project a new inertial loading system was developed at UNR to test single
cantilever-type columns on shake table under biaxial excitations. The aim of the test setup is
to have a supporting structure that carries safely the vertical component of the inertial mass
(superstructure weight) but allows transfer the inertial forces from the structure to thespecimen. A similar structure that allows dynamic excitation in one direction was developed
at UNR ten years ago (Laplace, 1999).
The new system is composed by a 3D four columns frame and a platform that sets on ballbearings located at the top of the columns. The platform is connected to the RCC specimen
through links in two perpendicular directions, which transfer shear and torsion but not axial
load (Fig. 3a). Additional mass is set on the platform to simulate the weight of a portion of
the bridge superstructure and this can be distributed in an asymmetric configuration to induce
torsion in the system. In addition, a safety system was designed to catch the platform in theevent of large displacements or specimen collapse.

56

Table 1: Material properties

Days

Concrete Compressive Strength [MPa]Circular

28

Interlocking

Footing

ColumnFooting

Column

33

28

36

27

Steel Properties

No.3

No.4

W2.9

W5.0

Yield stress [MPa]

423

448

400

400

Yield strain

0.00220.00230.00240.0024

Strain at hardening

0.012

0.0075N.A

N.A

Peak stress [MPa]

653

712

541

541

Strain at peak

0.124

0.115

0.115

0.126

Fracture stress [MPa]

561

687

537

484

Fracture strain

0.195

0.151

0.154

0.138

Figure 1: Geometric configuration and reinforcement for circular RCC*.

Figure 2: Geometric configuration and reinforcement for double interlocking RCC*.
* Unit conversion 1 in = 25.4 mm.

57

The axial load is applied directly to the specimen through a center-hole ram equipped with a
servo-valve. The ram is connected to the specimen throughout an unbonded prestressed bar
placed in an ungrouted conduit at the middle of the column and anchored at the footing. It is
important to note that the main purpose of the prestressed bar is to induce the required level ofaxial load in the columns rather than increases its displacement capacity as has been found inother studies (Sakai et al., 2006).
Since the designed system does not induce secondary moments (PD-effects) in the specimen
and the unbonded prestressed bar inside the column would generate restoring lateral forces,
additional dynamic actuators will be located at the top of the specimen to induce the equivalent
force to have PD effects and to compensate the restoring force throughout hybrid simulation(Fig. 3b).
In view of the complexity of the system in terms of the active control of dynamic actuators,
the test program was divided in two phases. At the beginning a set of two circular and twointerlocking columns will be tested without any axial load or PD effects. A second phase will
incorporate all the effects.

a: Without axial load.

b: With axial load (prestressed bar + actuators)

Figure 3 - Inertial loading system (Bidirectional Mass Rig).

Analytical Investigation
Analytical models were developed to anticipate the seismic performance of the specimens
and to determine the appropriate input loadings to be used during the tests. Time history inelastic
58

analysis have been performed using OpenSees (Mazzoni et al.,2006). Analytical models of
single cantilever-type columns with lumped mass as well as models of the specimens including
the inertial loading system were studied under different levels of earthquake excitations and mass
distribution to determine limit states in the behavior of the columns during the tests.
The biaxial flexural behavior of the columns was simulated using a distributed plasticity
model throughout uniaxial fiber elements (element inelastic beam-column in OpenSees). The
stress-strain properties of the unconfined and confined concrete were simulated using theMander’s model (Mander et al., 1988). For that, the actual strength of the concrete measured
from cylinders at 28 days was used. Likewise, the longitudinal reinforcing steel was idealizedusing the uniaxial steel material model developed by Chang and Mander (1994). The actual
stress-strain backbone curve measured from coupons was used as the input parameter for the
steel material model. Also, the reinforcement slippage was included in the models in the form of
additional rotation at the plastic hinge location.
Since inelastic fiber models for torsion are still under development (Mullapudi et al., 2008), a
reduction factor of 20% the elastic torsional stiffness (GJ) was used to take in account thetorsional cracking of the concrete in agreement with the Seismic Design Criteria (CALTRANS,
2006).
To estimate the lateral load and displacement capacities of the specimens moment-curvature
analysis were performed. Table 2 summarizes the capacities of the circular and interlocking
columns. Once the capacity was estimated, a series of nonlinear time history analysis wereconducted to select the input motion to be simulated in the shake table test. As was mentioned
before, five cases of mass distribution were studied to determine the largest torsional demand on
the specimen.
Table 2: Lateral load capacities of the specimens
Circular Columns P=0
Properties
Radial0.00034φy
My (kN-m)
177
0.00584φu
Mu (kN-m)
223
8.29
μΔ
Vu (kN)
122

Propertiesφy
My (kN-m)φu
Mu (kN-m)μΔ
Vu (kN)

1 in = 25.4 mm

Interlocking Columns P=0
Short dimension
Long dimension
0.00040.0003158
229
0.007420.00431
177
253
11.5
9.3
98
138

59

1 in = 25.4 mm

Ground motions
The two horizontal components of the 1940 Imperial Valley earthquake (El Centro), the 1994
Northridge earthquake, the 1992 Petrolia at Mendocino earthquake and the 1995 Hyogo-ken
Nanbu earthquake (Kobe) were used as the input motions. The earthquake records for Northridge
and Petrolia were scaled to have a hazard level of 2% of exceedence in 50 years (Zhang and Xu,
2008). The amplitude of the records was increased until the maximum capacity of the analyticalmodel was achieved. Also, the time axis of the input motions was compressed to account for the
specimen scale factor. Factors of 0.58 and 0.5 were used for circular and interlocking specimens,
respectively.
From the dynamic analysis, it was found that the record Petrolia at Mendocino and Sylmar at
Northridge amplified by factors of 2.0 and 1.8, will induce the maximum displacement ductility
demand on the circular and double interlocking specimens without exceeding the shake tablecapacity. The maximum accelerations imposed in both horizontal directions were 0.9g and 1.2g,
for Petrolia and 1.1g and 1.5g for Sylmar ground motion.
Column C1 – Experimental Results
The first circular specimen (C1) was tested under the two components of the Petrolia earthquake.
For this test no axial load was applied, and also the distribution of the mass was symmetric toinduce low torsion. The column was subjected to multiple motions, increasing the amplitude insubsequent runs. Small increments (10% to 20% of the real earthquake) were initially applied todetermine elastic properties and the effective yielding, after that, the amplitude of the records
was successively increased until failure. Signals of white noise were applied to the model tomeasure the chance in period and damping ratio between runs. Table 3 summarizes the measured
response of the column during the test.
Table 3: Measured Response of Specimen C1
Earthquake
0.1xPET
0.2xPET
0.4xPET
0.6xPET
0.8xPET
1.0xPET
1.2xPET
1.4xPET
1.6xPET
1.8xPET

MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN
MAX
MIN

Accel. in X
(g)
0.105
‐0.0860.191‐0.1850.363‐0.4440.455‐0.5880.590‐0.6670.692‐0.7370.806‐0.8080.887‐0.8680.981‐0.9391.098‐0.993

Accel. in Y
(g)
0.050
‐0.0490.106‐0.1040.255‐0.2380.337‐0.3810.466‐0.4560.607‐0.5700.752‐0.6560.863‐0.7621.039‐0.876
1.175‐0.965

Rel. disp. in X
(mm)

Rel disp. in Y
(mm)

10.83
‐10.33
13.55‐18.15‐20.3324.98‐20.6424.97
36.25‐29.4338.39‐97.0461.38‐47.5974.08‐56.22
86.97‐65.48‐59.48‐316.66

6.26
‐6.86
22.46
‐23.27‐11.67
9.81‐12.81
9.32
12.91‐17.6748.38‐15.7321.40‐28.6625.83‐34.4130.17‐40.09156.0032.81

60Force in X
(kN)
42.43
‐52.1868.83‐95.5899.45‐115.69
93.34‐103.29111.65‐82.56121.39‐65.64123.72‐67.80125.73‐66.53127.03‐67.50124.16‐71.81Force in Y
(kN)
24.72
‐25.9345.57‐40.6556.01‐57.9650.33‐49.7049.23‐58.1448.41‐66.4644.27‐70.4745.71‐73.4345.14‐73.9549.24‐74.40

The behavior of the specimen was controlled by the biaxial effect of bending, with horizontal
cracks distributed on the height of the specimen, as well as some inclined cracks at the plastic
hinge zone.
The first yielding was observed during 0.2xPET, after this motion, horizontal cracks spreadedout from the bottom of the column until the motion 1.0xPET, where the concrete cover at thesouth side spalled out. The maximum force was recorded at 1.6xPET, at this stage most of thetransverse reinforcement was exposed. Finally the failure of the column was observed at
1.8xPET, at this stage some of the longitudinal bars undergone buckling and large degradation of
the concrete core was observed. Figure 4 shows the final damage state of the column, while
figure 5 shows the displacement history and the hysteretic behavior of the specimen in
longitudinal direction (X). On the same plot is shown the predicted behavior using the analyticalmodel.

Figure 4 – Specimen C1 after failure

Figure 5– Displacement history and hysteresis for specimen C1

61

Concluding Remarks
The new inertial mass system to be used on bidirectional shake table tests at UNR represent asignificant advance in the simulation of single RCC under simultaneous loads induced by realtime earthquake motions. One of the most important characteristics of this system is that itallows the interaction between bending and torsion with or without axial load.
The analytical model used to simulate the seismic performance of the specimen and theinertial mass system as shown in Fig.4 predicted the answer reasonably well in terms of relative
displacement and hysteretic behavior.
Preliminary analytical and experimental results found at UNR and by researchers from otherinstitutions involved in the project have shown that the interaction between loads have asignificant effect in the capacity of reinforced concrete bridge columns under seismic loads.
These results are being used to develop analytical tools and new inelastic models for reinforced
concrete columns that in turn will assist in the development of new design methodologies.
Acknowledgements
The research presented in this paper was founded by The National Science Foundation underGrant No.EMS – 0530737. The assistance of Ian Buckle, Patrick Laplace, Chad Lyttle and thestaff of the Large Scale Structural Laboratory at University of Nevada Reno is gratefully
acknowledged.
References
ACI Committee 318 (2008), "Building Code Requirements for Structural Concrete (ACI 318-08)
and Commentary." American Concrete Institute, Farmington Hills, Michigan, USA, 465 pp.
California Department of Transportation. (2006) "Bridge Design Specifications." Engineering
Service Center, Earthquake Engineering Branch, California, USA, 250 pp.
Chang, G. and Mander, J. (1994). "Seismic Energy Based Fatigue Damage Analysis of Bridge
Columns: Part I – Evaluation of Seismic capacity." NCEER Technical Report 94-0006.
Correal, J.F., Saiidi, M.S., and Sanders, D.H. (2004). "Seismic Performance of RC Bridge
Columns Reinforced with Interlocking Spirals." Report No. CCEER-04-6, Center for Civil
Engineering Earthquake Research, Department of Civil Engineering, University of NevadaReno, USA, 438 pp.
Zhang, J. and Xu, S.Y. (2008). " Seismic Response Simulations of Bridges Considering ShearFlexural Interaction of Columns. " Proceedings of the 6th National Seismic Conference on
Bridges & Highways, Paper No. P29, Charleston, South Carolina, USA.
Mander J.B., Priestley, M.J.N. and Park, R. (1988). "Theoretical Stress-Strain Model for
Confined Concrete." Journal of Structural Eng. 114 (8), pp 1804-1826.

62

Laplace, P., Sanders, D.H., and Saiidi, M.S. (1999). "Shake Table Testing of Flexure DominatedReinforced Concrete Bridge Columns." Report No. CCEER-99-13, Center for Civil Engineering
Earthquake Research, Department of Civil Engineering, University of Nevada-Reno, USA, 438pp.
Mazzoni, S., McKenna, F., Scott, M., and Fenves, G.L. "Open System for Earthquake
Engineering Simulation (OpenSees), 2006." User manual, version 1.7.3, Pacific Earthquake
Engineering Research Center, University of California, Berkeley,California,USA,
http://opensees.berkeley.edu/OpenSees/manuals/usermanual/index.html.
Mullapudi, T.R., Ayoub, S. and Belarbi, A. (2008). "Effect of Coupled Shear-Bending
Deformations on the Behavior of RC Highway Structures Subjected to Extreme Seismic
Loading." Proceedings of the 6th National Seismic Conference on Bridges & Highways, Paper
No. P19, Charleston, South Carolina, USA.
Sakai, J., Jeong H. and Mahin, S.A. (2006). "Reinforced Concrete Bridge Columns that ReCenter Following Earthquakes." Proceedings of the 8th U.S. National Conference on Earthquake
Engineering, Paper No. 1421, San Francisco, CA.

63

blank25th US-Japan
Bridge Engineering Workshop

Session2

Inspection and Management
Analysis of Periodic Inspection Results of Highway Bridges in Japan
By Koichi Ikuta, Takashi Tamakoshi, and Masanori Okubo
Bridge Inspection and Management in California
By Barton Newton
Measures for Strategic Preventive Bridge Management of Tokyo Metropolitan Government
By Taro Awamoto and Sentaro Takagi
Instrumentation and Monitoring of I35W St. > Anthony Falls Bridge
By Catherine French, Carol Shield,Henryk Stolarski,Brock Hedegaard,
and Ben Jilk
Study for Bridge Renewal and Repair by Osaka Municipal Government
By Shinsuke Yumoto, Tetsuya Yokota, and Yasutomo Komatsu

Page Left Blank

ANALYSIS OF PERIODIC INSPECTION RESULTS OF
HIGHWAY BRIDGES IN JAPANTakashi TAMAKOSHI1, Masanori OKUBO2, Koichi IKUTA3
Abstract
Analyses of initial inspection results, relationship between age and damage,progress of damage, and fatigue damage were conducted using periodic inspection
results as a part of the research on rational periodic inspection method to grasp
conditions of nationwide bridges from the uniform point of view efficiently.
Background
In Japan, under limited budget and stuff according to social and economical changethat low fertility and aging are in progress, to maintain service level of its huge road
network strategically and efficiently is required. For this, prevention maintenance thatcountermeasures are conducted at the early stage before the abnormalities become
serious, not the countermeasures like symptomatic treatment for remarkable damages,
is needed, and strategic and rational management that contributes on reducing life
cycle cost for long time considering whole road asset is also needed. Furthermore, itcomes important to maintain roads over the nation as network.
Under this background, paying attention to bridge inspection which is the mostbasic action to get information needed for bridge maintenance, analyses of periodic
inspection results were conducted as a part of the research about the rational periodic
inspection method to grasp conditions of nationwide bridges from an uniform point ofview efficiently.
State of bridge inspection in Japan
Bridges operated by MILT are inspected with periodic inspection manual, which
was revised on March 2004 from that edited on July 1988 to conduct inspection more
efficiently with the new findings in maintenance and inspection results.
1) Frequency of inspection
For inspection frequency, initial inspection should be done within 2 years after thebridge is opened to the traffic, after that the bridge should be inspected every 5 years.
1 Head, Bridge and Structures Division, Road Department, NILIM, Japan
2 Senior Researcher, Bridge and Structures Division, Road Department, NILIM, Japan
3 Researcher, Bridge and Structures Division, Road Department, NILIM, Japan

67

Initial inspection is aimed to increase durability and to reduce life cycle cost by
taking measure at early stage against initial defects which appear within several years
after opening to the traffic. The initial inspection results are meaningful as the initial
data for various maintenance works like repair and reinforce within the service life of
the bridge.
2) Method of periodic inspection
The method of periodic inspection has basically followed the precedent: the method
to evaluate every member considering every usual kinds of damage by close visualinspection.
This is the result of the judge that this inspection method is the most rational andindispensable considering economy at this point to detect partial damage and to grasp
the progress of serious damage which develops extremely rapidly at some parts in the
point of view of keeping structural safety of the bridges, preventing damage for a third
person, realizing preventive maintenance by detecting the indication of damage at earlystage. That is, considering the limit of stuff and budget, there are some places to
examine to use distant view or to introduce advanced technical skills likenon-destructive inspection methods, but many serious damages to affect the safety ofbridges like short of sectional area in steel members by crack or corrosion are difficult to
detect by distant view, and the rational methods to cover the risk of missing thosedamages have not been established.
3) Evaluation of inspection results
Two types of evaluation of inspection results were introduced when the periodic
inspection manual was revised on 2004: “evaluation of degree of damage”- it is a recordof facts of degree of damages, and “judgment of countermeasure”- it is a first diagnosis
about function of bridge and structure by proper engineers.
Information which is grasped by inspection like damages and aging is not only
indispensable to inferring cause and to evaluating performance, but also used for future
estimation or trend analysis. Information should be objective data based on an uniform
standard.
On the other hand, evaluation and diagnosis by engineer with technical knowledge
accompanied with inspection are dispensable to various damages and their effects’ being
measured. It is very important for administrator to decision-make properly about needfor traffic regulation and countermeasure like repair and rehabilitation gasping theperformance of bridges from inspection. For this, it is important to get not only facts of68kind of each damage and degree of their development, but also damage’s effect tobridge’s performance and view about response that should be done from the damage’s
effect. For example, some fatigue cracks of steel members develop rapidly to dangerous
state of the bridge even if the crack is small when it is detected.
For “evaluation of degree of damage”, every
parts is evaluated in the divided unit of each

Table 1. Classification of degreeof damage

member because the evaluation is the basic data

Evaluation

for investigation for rationalization of futuredivision

inspection.

Damages

detectedatperiodicａinspection are evaluated at every element and atｂevery degree of damage, and the classificationｃmeans objective facts indicating the degree ofｄdamage. The classification of degree of damage isｅDegree of damage
Small

Large

shown in Table.1. Damages are evaluated to theclassification of “a”, “b”, “c”, “d”, and “e” by degree of damage, and “a” is the least seriousand “e” is the most serious.
For “judgment of countermeasure”, on the other hand, the unit of member withsome sizes, like main girder or pier, is evaluated because“judgment ofcountermeasure” is qualitative evaluation which can be called an advice foradministrator about countermeasure in relation to performance of the bridge. The
classification of evaluation is shown in Table.2.
Table. 2 The classification of evaluation at judgment of countermeasure
Evaluation

Content of evaluation

divisionＡ

Damage is not confirmed or little, so repair is dispensable.

Ｂ

Repair is needed

Ｃ

Repair is needed immediately.

Ｅ１

Emergency response is needed for structure safety.

Ｅ２

Emergency response is needed for other reason.

Ｍ

Response through maintenance construction work is

according to the situation.

necessary.
Ｓ

Detailed investigation is necessary.

Analysis of periodic inspection results
Various kinds of analysis of the periodic inspection results were conducted for
improvement of periodic inspection because one round of periodic inspection had almost

69

finished.
1) Analysis of initial inspection results
Damages were detected in most of bridges in first inspection. The number of bridges
as the objects of first inspection was 198, which were within 2 years after opening to thetraffic and after construction completed, and damages were detected in 193 bridges(97%). The most detected damage was deterioration of rust-proof performance andcorrosion in steel members and crack, leakage and isolated lime in concrete members.
Abnormality of anchorage zone of PC member and abnormality of performance ofbearing were detected as the serious damage to affect performance of bridge.

Fig.1 Initial inspection results (main members)2) Relationship between age and damage
As a whole, the rate of abnormality
seems to increases in accordance with age.
Fig.2 shows states of damage in periodic
inspection results at age after opening tothetraffic.

Bridges’deterioration

characteristic has huge variation because
it depends on bridge’s structure andenvironment, and it can be inferred that
degree of deterioration and damage of eachbridge differ. However, as a whole, the
trend

that

the

rate

of

abnormality

increases in accordance with age is notable.

70

Fig.2 The rate of bridges in classificationof measure3) Analysis about progress of damages①

Crack at concrete members (part)
Among the relationships between age and degree of various kinds of damage, theexample of crack at concrete main girder is shown in Fig.3, and the example ofabutment is shown in Fig.4. The transitions of main girder and abutment are
different. On the other hand, it can be inferred that when crack generated even atmain girder, the crack certainly progresses within 5 years.

Fig.3 Relationship between age and degree of damage(crack at concrete main girder)② Crack at concrete members (environment)
The transition of two periodic inspection
results was analyzed classifying chloride
damage area and other area to grasp theeffect of environment regardless age. The
result is shown in Fig.5. Marcov transition
probability based on Fig.5 is shown in Fig.6.
It is recognized that the deterioration speed
in is a little faster chloride damage area whenFig.4 Relationship between
age and degree of damage(crack at concrete abutment)

degree of damage is light.

Fig. 5 Transition probability of generation of crack at concrete members

71

Fig.6 Marcov transition probability of crack at concrete main girders③ Crack at concrete members(structural type, crack pattern)
Age and state of crack generation at PC post-tensioned hollow slab bridge and PC
post-tensioned T section girder bridge are shown in Fig.7 and Fig.8.

Fig.8 Age and state of crack generation(PC post-tensioned T section girder bridge)

Fig.7 Age and state of crack generation(PC post-tensioned hollow slab bridge)72For PC post-tensioned T section girders, age and state of crack generation in eachpattern whose generation rate is high is shown in from Fig.9 to Fig.12

Fig.9 Number of damaged span in each damaged pattern(PC post-tensioned T section girder bridge)

Fig.10 Age of PC post-tensioned
T-section girder bridges and stateof crack generation (Pattern①)

Fig.11 Age of PC post-tensioned
T-section girder bridges and stateof crack generation (Pattern②)

Fig.12 Age of PC post-tensioned
T-section girder bridges and stateof crack generation (Pattern④)73

④ Corrosion at steel girder (differenceof material)
Transition

probability

about

corrosion at steel girder also calculated
atgeneralpainting

systemandheavy-duty painting system. The resultis shown in Fig.13.
The trend is that the speed ofdevelopment
systemisatgeneralquickerthanpainting
thatofheavy-duty painting system for small

Fig.13 transition probability of corrosion
at steel main girders

damage, but the difference is notnotable for large damage.
4) Analysis about fatigue
In the crack of steel member, the relationship of fatigue damage, age, andcumulative traffic volume was analyzed. The results of steel plate girder bridges andsteel box girder bridges are shown in Fig.14, and the results of steel truss bridge areshown in Fig.15. The trend is recognized that the rate of generation of crack increases
when the age gets more than 30 years, and cumulative large vehicle volume gets 250
thousand vehicles・year in steel plate girder bridges and steel box girder bridges, andthat the rate of generation of crack increases when the age gets more than 30 years insteel truss bridges.

Fig.14 Rate of number of span that crack generates (Steel plate and box girder bridges)

74

Fig.15 Rate of number of span that crack generates (Steel plate and box girder bridges)
The results about RC slab are shown in Fig.16. It could not be seen the damage’s
significant relationship with age and traffic volume. The damages might include thedamage except for fatigue.
For fatigue damage of RC slab, damage progress patterns are being analyzed
including states of leakage and isolated lime with crack. The result of analysis of crack
patterns is shown in Fig.17

Fig.16 Rate of number of span that crack generates (RC slab)

75

Fig.17 Analysis of crack patterns of RC slab
5) The relationship of degree of damage and classification of measure
The relationship of degree of damage and classification of measure in the case ofcrack at concrete main girder is shown in Fig.18. Both evaluations are not coincident. It
can be inferred that engineers evaluate in classification of measure considering thebridge’s unique condition like environment.

Fig.18 The relationship of degree of damage and classification of measure(crack at concrete main girder)76

Summary
Periodic inspect ion data has been analyzed to rationalize periodic inspection to
get basic data in maintenance strategy and to grasp state of bridges for prevention of
emergent accident and timely countermeasure like preventive maintenance.
The followings can be mentioned as findings from this analysis:
1) It is highly possible that there is a limitation to model deterioration characteristic
because progress of deterioration is related to various factors such as construction
environment, traffic condition, structural characteristic of whole bridge and every
member.
2) Therefore, it can be referred that inspection method and frequency should be
optimized according to not only common conditions such as structural characteristic
and traffic but also individual part or member of the same bridge.
3) Degree of damage as “Objective fact” used for deterioration prediction and statisticalanalysis not always corresponds with evaluation based on performance of the bridge.
Both of them are needed for maintenance.
4) Evaluation based on performance of the bridge should be considered each viewpointof damage risk, emergency, preventive maintenance and so on for propercountermeasure.
Analyses of inspection results will be continued for an establishment of therational bridge inspection system.

77

blankBRIDGE INSPECTION & MANAGEMENT IN CALIFORNIA
Barton J. Newton, PE1
Abstract
This paper provides an overview of the California Department ofTransportation (Caltrans) Bridge Management Program including inspection andinvestigations; special investigations including NDE/NDT technology; data
management; prioritization and programming decision making; accomplishing bridge
maintenance work; rehabilitation and replacement programs; and measuringperformance.
Introduction
California – the Golden State - boasts a growing, diverse population of 38
million people. For more than 50 years, California has embodied a spirit of innovationand entrepreneurship that has given birth to new technologies and industries whilebuilding the eighth largest economy in the world.
The backbone of the economic miracle that is California has been theconstruction, operation and maintenance of the finest highway system in the world.
Caltrans is helping propel California’s economic engine by spearheading $9.5 billion
worth of construction work that is supporting more than 170,000 jobs in this state.
Today, that system of highways and bridges safely and reliably carries
millions of people and dependably delivers commerce to and from every corner of thestate, the nation and the world. Not only is the work we are doing in California
important to economic recovery, it’s critical to America’s economic strength in theworldwide economy as products that flow into and out of this country rely on a fullyfunctional nationwide transportation system. What we all do in the transportation
business is important.
The California Department of Transportation is responsible for maintaining
more than 15,000 centerline miles (50,000 lanes miles) of pavement and 12,600
bridges on the state highway system.
Taking care of its bridge inventory has been a critical part of California’s
success in making sure its transportation program continues to deliver reliable service
year after year. In 1927, state leaders initiated a study to determine and document thecondition, safe load carrying limits and other restrictions of each public bridge in thestate and to develop progress reports with work recommendations. Out of that effort
was born California’s bridge inspection and maintenance program.
___________________________________
1
State Bridge Maintenance Engineer, State of California Department ofTransportation (Caltrans)79

Inventory
The California Department of Transportation, as the owner and operator of thestate highway system, is responsible for the protection and preservation of the state’s
bridge inventory which has an estimated replacement value of more than $52 billion.
California’s bridge inventory is aging with about two-thirds of structures on
the state highway system having been built between 1950 and 1980. Approximately
92% of California’s state highway bridges are concrete, 7% steel with timber andcomposite materials comprising the remainder. Concrete box girders are the mostcommon bridge type.
Bridge Program Goals
California’s bridge inspection and maintenance program has five goals:
1. To ensure the safety of the traveling public as required by Federal and StateLaw.
2. To manage the bridge assets to an agreed level of service.
3. To collect inventory and condition data
4. To recommend needed repairs, rehabilitation and/ or replacement
5. To determine safe load carrying capacity of each structure.
Structure Maintenance & Investigations
As part of the Caltrans Division of Maintenance, the Structure Maintenance &Investigations (SM&I) organization is charged with ensuring the structural integrity of
the 12,600 state highway bridge inventory and 12,500 bridges owned by local
government agencies.
The SM&I staff performs bridge inspections and engineering investigations inaccordance with federal regulations, makes repair work recommendations;investigates hydraulic problems and scour potential; determines the safe load carrying
capacity of all bridges; reviews and approves encroachment permits and air spacelease proposals for state-owned structures; manages the State’s bridge funding
programs; delivers plans specifications and estimates for maintenance andrehabilitation on state-owned bridge projects; and coordinates the protective coating
work on more than 800 steel state-owned highway bridges. SM&I is also the
Department lead responsible for responding to any type of bridge related emergency
including earthquakes, floods, over height vehicles, and tanker fires.
There are over 180 bridge engineers, structural steel technicians and support
staff in six offices within SM&I. Those offices are listed below.
1. The Executive Office located in Sacramento.
2. The Office of Structure Investigations, North located in Sacramento is
responsible for performing bridge inspections of all types on State andLocally-owned structures in the upper 49 counties of the State.
3. The Office of Structure Investigations, South located in Los Angeles is
responsible for performing bridge inspections of all types on State and80Locally-owned structures in the lower 9 counties of the State andmaintenance repair design of state owned bridges in the southern partof the State.
4. The Office of Toll Bridge Investigations located in Oakland is
responsible for the inspection, maintenance repair design and generalcustodianship of the major toll bridges in San Francisco bay area.
5. The Office Specialty Investigations and Bridge Management located in
Sacramento is responsible for the special Fracture Critical andUnderwater Investigations for specified bridges statewide. It also
manages the Statewide Bridge Preservation Program (Capital Funded)and the Statewide Bridge Maintenance Program, and is responsible forthe data management for all bridges in the state.
6. The Office of Structural Design and Analysis is located in Sacramento
and provides design services for bridge repair projects, establishes thesafe load carrying capacity for all bridges and reviews encroachment
and transportation permits. In addition, hydrologic and hydraulicevaluations and investigations are also performed for all existingbridges over waterways by this office.
Inspection
The foundation of the bridge inspection program is grounded on therequirements of the National Bridge Inspection Standards (US Code of Federal
Regulations 23 CFR 650). California uses element level inspection protocols, visual andnon-destructive testing techniques to assess the condition of each bridge on regularintervals between two to five years. The interval is based on the minimum federal
requirements as well as a risk assessment of each bridge that considers type, actual
condition, and redundancy of the structure and the judgment of the inspector.
Ensuring the safety and reliability of California’s bridges is achieved throughan ongoing effort of routine, fracture critical and underwater inspections performed bylicensed engineers and specially trained structural steel inspectors. Federal regulations
require that routine inspections be conducted every two to four years, fracture criticalinspections every two years and underwater inspections every five years.
Routine Inspections
During routine inspections, inspectors who are licensed engineers look for any
signs of distress that could compromise the structural integrity of a bridge. These
inspectors assess the structural condition of the bridge, looking for any indications ofpotential problems. They examine the bridge deck – the riding surface- looking forcracks or other signs of distress. The inspectors, drawing on their expertise, can tell if
cracks are superficial or signal something more serious. The conditions are
documented, monitored, and repairs recommended if necessary. Inspectors also mayorder additional investigation if needed, such as NDE techniques or taking coresamples of the concrete deck in for testing to the Caltrans laboratory to document anddetermine the extent of cracking etc.
The same process is followed on the superstructure—the structural membersthat support the deck. The Inspectors, relying on their knowledge and expertise and81using their trained eyes to identify any signs of distress. They will climb piers anduse sophisticated equipment to access structural members of the bridge. Again,
conditions are documented and repairs recommended as needed.
Finally, inspectors examine the surround areas and the substructure—thefoundation of the bridge—looking for any signs of deterioration. As with the deckand superstructure, conditions of the substructure are documented and repairs
recommended in a bridge inspection report.
Fracture Critical Inspections
A nine member team of engineers and structural steel inspectors is responsiblefor inspecting more than 1,000 state and local agency steel structures. A fracture criticalbridge has a steel member whose failure could cause a portion of or the entire bridge tocollapse. The I-35 West Bridge that collapsed in 2007 in Minnesota was a fracture
critical bridge.
All areas of fracture critical bridges are required to be inspected visually, within arms
reach, every 2 years with the aid of lift equipment or an under bridge inspection truck toplace the inspector within arms reach of the fracture critical element. Many times
fracture critical inspections utilize non-destructive testing equipment to help find cracks
in critical members that can be invisible to the naked eye.
Underwater Inspections
Caltrans employs a full commercial certified underwater inspection team to
perform bridge inspections and to respond to emergencies such as ship collisions withbridges.
SM&I’s 11-member Underwater Inspection Team is responsible for ensuring thestructural integrity of the supporting piers for more than 530 bridges including major bay
crossings like the San Francisco-Oakland Bay, the Richmond-San Rafael and the San
Diego-Coronado bridges. Inspectors look for any damage and scouring of bridge piers.
Underwater inspections of bridge piers are conducted in waterways from the
Pacific Ocean and the state’s major rivers to the California Aqueduct. These underwaterinspections are designed to detect any loss of strength or the potential of scouring on
bridge piers below the water surface.
Non-Destructive Testing
Safety is enhanced through these inspections and by “rating” bridge components,
such as the deck, superstructure, and substructure, and by the use of non-destructive
evaluation (NDE) methods and other advanced technologies. Visual inspection is theprimary method and technique used to perform bridge inspections in California. Type,
location, accessibility, and condition of a bridge, as well as type of inspection, are someof the factors that determine what methods of inspection practices are used. When
problems are detected, or during the inspection of critical areas, more advanced NDE
tools are employed. On occasion, destructive tests are conducted to evaluate specificareas or materials of concern, or to help identify appropriate corrective work.

82

The inspection team utilize many NDE methods for steel structures such as
liquid penetrant testing (PT), ultrasonics (UT), eddy current (EC), radiography (RT) andmagnetic particle testing, strain gauging, acoustic emission (AE), and x-ray technology.
Methods utilized during inspections of concrete structures include impact echo,
infrared thermography, ground penetrating radar, and strain gauges for concrete
structures and elements.
Load Ratings and Transporation Permits
The Bridge Ratings and Analysis Branch are responsible for establishing the safe
load carrying capacity for all bridges and reviewing all overweight transport permits.
The unit completed 126 load rating analyses during the 2008/09 Fiscal Year,reviewed nearly 200 extra legal truck loads weighing more than 250,000 pounds. Morethan a third of these loads were more than 800,000 pounds. They also successfullyrouted one super load of more than one million pounds and four super loads of nearly
one million pounds. All five loads were transformers.
The four loads of nearly one million pounds were manufactured in Southeast
Asia, shipped to the Port of San Diego and transported to a storage facility in El Centro.
The transformer weighting more than one million pounds was brought by rail from
Houston to Fontana in San Bernardino County. The transformer was transported over
California’s road network to the Mexico/California border for eventual delivery to a new
natural gas power plant in Rosarito, Mexico that will supply power to Tijuana and San
Diego.
Bridge Data and Management
SM&I serves as the repository of data and information about every state andlocal agency bridge in the state. This information includes a historical record of thecondition of each bridge on the state highway and local road networks, including all the
as-built engineering plans and all completed bridge inspection reports. Bridge inspection
reports are prepared following each inspection. These inspection reports provide a living
history of each structure. Our library of information contains more than on million
documents and gives engineers easy access to the entire structural history of each bridgein California
Bridge needs arise from the deterioration of bridge components, vulnerabilities toscour and seismic forces, changing safety standards and emergencies/accidents involving
bridges.
Bridge Inspection Reports serve as the basis for initiating repairs in a timely andcost effective manner to ensure that each bridge is safe and well maintained.
Caltrans uses the AASHTOWare Pontis Bridge Management System and othertools to help evaluate the needs and prioritize repairs to maximize the benefit of available
preservation dollars. Caltrans employs a life cycle cost analysis approach to compare
alternatives at the project level.

83Caltrans has recently begun a movement to a utility based benefit cost analysis
approach that can combine physical condition and risk into a single utility that can beobjectively compared across bridges.
Bridge Maintenance Design
There are four maintenance design units in SM&I, two in Sacramento and one
each in Los Angeles and Oakland. A total of 32 design engineers and structural design
technicians are responsible for delivering plans, specifications and estimates for a variety
of bridge deck, superstructure and substructure structure preservation and repair projects.
The projects include methacrylate deck treatments and polyester overlays, joint
seals, bridge rail repairs and foundation protection work. During the 2008/09 fiscal year,
SM&I designers delivered engineering plans for more than $135 million to repair 983
bridges around the state.
Research
SM&I is actively involved in advancing the science of engineering to extend theservice life and reliability of the state’s bridges. In conjunction with the CaltransDivision of Research and Innovation, SM&I is involved in a variety of structuralresearch efforts including the effectiveness of deck rehabilitation strategies and the useof sonar imaging to validate the structural integrity of bridge piers.
Preserving the Bridge Inventory
California has an ambitious bridge preservation program designed to extend thereliable, productive service of the state’s bridges. The preservation effort is comprisedof three main components: Caltrans inspection and maintenance crews, major
maintenance contract projects and the State Highway Operation and Protection Program(SHOPP).
Work recommendations are generated from information collected through bridge
inspections. The work recommendations are addressed by the bridge crews orprogrammed into projects to fix joint seals, methacrylate deck crack sealing, bearing
replacements, painting, polyester concrete overlays and replacing approach slabs.
Projects are prioritized to address the most pressing needs first to ensure
continued reliability of the system. Peer reviews are conducted by SM&I engineering
staff on all bridge replacement and major repair project exceeding $4 million to makesure that the recommended repair strategy is the best engineering and most cost effective
alternative.
Caltrans invests $67 million annually for inspection, bridge maintenance andbridge painting crews. The inspection effort is augmented by bridge maintenance crews
performing minor repairs such as filling potholes on bridge decks and repairing minor
bridge rail damage. They also respond to emergencies such as filling holes in a bridge
deck to restore traffic service on a temporary basis until a permanent repair can becompleted.

84

Bridge Paint crews are responsible for painting more than 800 steel bridges
around the state. There are permanent painting crews assigned to some of the large Bay
crossings such as the San Francisco-Bay Bridge. There also are general crews
responsible for painting the other steel bridges up and down the state.
Caltrans is investing and additional $93 million annually for major maintenancecontract projects. These projects are designed to protect the bridges and extend theirservice life. Approximately 23% of California’s bridge inventory, 2,844 bridges, have
major maintenance repair work recommendations pending.
Bridges like all physical assets will deteriorate over time and require maintenance,
rehabilitation or replacement. The graphic below depicts the rate of progression ofbridges from good (green) to fair (yellow) to poor condition (red).
Bridge Crew

Maintenance
300 Bridges/Yr

9,400 Bridges

Rehab
40 Bridges/Yr

2,200 Bridges

1064 Bridges

The wear and tear of highway traffic, weather, and movement cause approximately 300
bridges to change from good to fair condition and another 40 bridges change fromneeding maintenance work to needing rehabilitation.
Caltrans bridge preservation approach is a three prong attack that strives tocounteract deterioration:
Maintenance Program
Capital Program
HM3
1. Bridge crews employed by Caltrans address minor preservation very(SHOPP)
quickly tokeep bridges in the green.
2. Major maintenance contracts are required when the scope of the work exceeds
what the crews can do.
3. Major maintenance contracts are designed to delay or prevent the progression
into the costly red rehabilitation/replacement area.
Major rehabilitation or replacement is required to address the bridges in the red area.
Measuring performance
The United States government uses inspection data compiled by Caltransengineers to calculate a complex formula called the sufficiency rating to determine
federal bridge program funding eligibility. The sufficiency rating combines thecondition and functional adequacy data collected on every bridge into a single aggregate
number. Sufficiency rating values range from 0 (low) to 100 (high). If the sufficiency
rating on a bridge is 50 or less and it is designated as “structurally deficient” or“functionally obsolete” the bridge qualifies for federal replacement funding. A low
sufficiency rating number does not necessarily mean that the bridge is in need of repair.

85

Caltrans identifies prudent repair and rehabilitation projects taking into
consideration many factors that may not be fully captured by these federal designations
and ratings. For example, none of the federal designations or ratings take into
consideration the need for seismic strengthening. Caltrans relies on therecommendations from its field assessments and analyses to develop repair projects.
These repairs may be minor and assigned to Caltrans maintenance crews or they may be
more involved and require further analysis or planning studies to determine theappropriate course of action. Caltrans utilizes bridge management system software toolsthat can forecast deterioration and conduct life cycle cost analyses along with multidisciplinary peer reviews to develop our ultimate projects. Project decisions are made ona bridge-by-bridge basis.
Crews are measured on their ability to retire identified work within specified
time frames. The Caltrans Five Year Maintenance Plan tracks the number of bridges
requiring maintenance contracts and the SHOPP measures the number of distressedbridges with rehabilitation or replacement needs.
Emergency response
Structures Maintenance and Investigations (SM&I) is the lead engineering
responder to all emergencies involving existing state highway structures.
Operating from its three offices: Sacramento (North), Oakland (Toll Bridges),and Los Angeles (South), SM&I engineers are responsible for all matters involving thestructural integrity of any highway structure damaged in natural disasters, accidents suchas high load hits or other incidents. In the event of a disaster, the official damage reportis prepared, stamped and signed by a licensed civil engineer at SM&I. The damage
report is included in the Caltrans Bridge management database (SMART), which ismaintained by SM&I.
Damage information is shared with the California Office of Emergency Services,which has responsibility for directing the statewide response to all disasters in theGolden State and the Caltrans Division of Maintenance which coordinates the responseof maintenance forces around the state.
SM&I engineers have been at the forefront in responding to some of the mostnoteworthy natural disasters in the past 20 years including the 1989 Loma Prieta and1994 Northridge earthquakes, the 2008 collapse of the MacArthur Maze connector ramp
in Oakland due to a tanker truck fire and the collision of an oil tanker into the protective
sheathing around one of the San Francisco-Oakland Bay Bridge’s piers that resulted in
one of the worst environmental catastrophes in state history.

86MEASURES FOR STRATEGIC PREVENTIVE BRIDGE MANAGEMENTOF TOKYO METROPOLITAN GOVERNMENT＊Taro Awamoto 1
Sentaro Takagi 2
Abstract
Tokyo Metropolitan Government (TMG), Bureau of Construction has done
periodic bridge inspection and has been corrective bridge maintenance. As aging
bridge is increase, the possibility that serious accident is high. Finance limitation
demand the peak project cost cut and reduce of total cost. Therefore we changed
maintenance management strategy from corrective to preventive. We adopted New
Public Management type Asset management. We developed some mechanism, which
is deterioration model, social benefit, project prioritization method, and computer
systems. Finally we planned “Long-Term Bridge Management Plan”. This paper
reports our measures for strategic preventive bridge management of TMG.
Introduction
TMG, Bureau of Construction, manages approximately 1,250 road bridges.
Many bridges constructed in two periods. First period is reconstruction period fromGreat Kanto earthquake damage on 1923.Second period is high economic growth
period from middle 1950s to early 1970s. (figure 1). For number of bridge, 52 percent
are Reinforced concrete bridges or Prestressed concrete bridges, and 40 percent are
Steel bridges. Whereas for area, only 12 percent is Reinforced concrete or Prestressed
concrete and 56 percent is Steel (figure 2).
Bureau of Construction developed periodic bridge inspection every five years
from 1987. The method is direct sight inspection as much as possible. Inspection
contents are thirty one, which is corrosion, fatigue of steel girder, crack of concrete,and so on. Bridge health index is decided two steps. First, the condition of bridge isgraded five ranks from ‘a’ to ‘e’ for member unit. Second, total bridge health index isgraded five ranks from ‘A’ to ‘E’ for span unit and bridge unit, with used two index
durability and safety. The deterioration of bridge progressed certainly, by the result ofbridge inspection from 1987 to 2002. The rank C and D, which is required torehabilitation, is increase (figure 3).The cause of steel girder bridge damage anddeterioration is below. First is corrosion and deterioration of painting. Second isdeformation caused by bearing has not work normally or foundation have moved.
Third is fatigue damage by repeated heavy load. The damages of concrete bridge areconfirmed fall of filling deck slab, free lime, and crack.
Bridge maintenance was corrective, when damage and deterioration was found,bridge took appropriate measures. TMG constructed many bridges in the 1960s, higheconomic growth period. We envisage the peak of bridge replacement cost is early2030s, so that huge budget is need. Supposed from present TMG budget, increase of
1
2

Senior Staff Member, Bureau of Construction, Tokyo Metropolitan Government
Profession Director, Bureau of Construction, Tokyo Metropolitan Government87

bridge management budget is not much expected. We need to try to decrease peak costof bridge replacement and total bridge management cost. As bridge is aging, damage
and deterioration is rapidly extended, it is difficult to the inspector find damage anddeterioration. Thus, in the case of too late finding damage and deterioration to practicecorrective measure, serious accident can occur. The I-35W bridge collapse in
Minneapolis, Minnesota in the United States, 2007. In Japan, steel truss diagonal
member broke on two bridges, Kisogawa Ohashi Bridge on National Road 23 andHonjo-Ohashi Bridge on National Road 7. Existent bridge is lack of usability andsafety. It is not satisfy present design standard. Remaining this condition, thepossibility of fatigue damage is high. A surrounded environment changed with
progress of technique. Asset management methodology developed and tried to
introduce on business in many countries. As the result of high level study, thetechnique of new material and new method of construction developed. Therefore themeasure to long-life bridge without block traffic flow, is enable.
On a basis of the change of these situations, TMG change management strategy
to preventive maintenance with asset management methodology. We decide try to
decrease peak cost of bridge replacement, to reduction total bridge management cost,
and to keep safety traffic. For important bridges, replacement of bridge is not executed
because the replacement cost of these bridges is much cost needed. To satisfy present
design and to make long-life bridges, maintenance cost decrease by performance based
design. This paper reports present condition of measures of TMG.
Asset Management
Asset management means generally the businesses which manage property
substitute for owner and investor. The term used to use securities firm and real estateagent.
Recently, in the United States and other countries, Asset management
methodology was introduced to operation and maintenance of infrastructure. FHWA
created an office of Asset management in 1999. Now, Asset management is acceptedamong many agencies in the United States. In Japan, some local government are
working on introduce of Asset management.
Japanese Asset Management method is two types. One type is LCC type, to try
minimum life cycle cost and another type is New Public Management (NPM) type, totry maximum difference between value and cost (JSCE(2005)). NPM type is theadvanced LCC type, to progressively and for invests. As we know, in Japan, local
governments except TMG have introduced LCC type Asset management. Only TMG
is introduced NPM type Asset management.
New Public Management is the movement to achieve public efficiency andprogress of quality of work, by widely introduced private enterprise management style.
The goal of the NPM type Asset management of TMG is, the government as agent ofTokyo metropolitan citizens, uses their tax efficiency to road infrastructure, and return
effective and efficient public service to citizen or user. More specifically, the result of
public investment, entrusted use of citizen tax, to maximum the effect tax payer
received. Bureau of Construction make investment project plan to difference between
benefit occurred from bridge exist and project cost.

88Deterioration Model
On preventive maintenance management, evaluate future condition of eachbridge by deterioration forecasting and make plan which can execute appropriate
measure on best timing is need.
We set original deterioration curve from past twenty decades bridge inspection
data collection, calculated last bridge life length. The deterioration curve is calculated
life-cycle analysis and non linear regression analysis. In the beginnings, we tried to set
standard deterioration curve, with use linear curve as general deterioration curve
calculation method. But there are almost no data the most danger health index ’E’, we
judged this linear curve do not mirror real deterioration and changed our policy to theabove method.
The process of deterioration curve set is below. First, we divided bridge
inspection data into twenty types by combination of three elements, member, material,and damage. And calculate the time to reach next rank by life-cycle analysis. Second,
we made four alternative curve types, exponential convex curve type, linear curve type,second polynomial curve type, and third concave curve type. Each bridge deterioration
curve is decided by select the most nearly curve type compared with the result of thelife-cycle analysis(figure 4).
Social Benefit
The achievement of NPM type Asset management of TMG is to maximize thedifference between benefit Tokyo citizens receive and project cost. The benefit isimaginary profit, and we call social benefit it. Cost benefit analysis, which is used ineconomic and civil field, supply that the measurement method of some benefit. Settingon each benefit contents, we focused that the variations in traffic flow with project
operation. We reasoned that the benefit Tokyo people received is difference benefit ofWith: the project is operated, and of Without: the project is not operated. We set fivebenefit contents, drive time shortening, driving cost reduction, comfort driving,environment load reduction, and traffic accident reduction.
Drive time shortening is based on the Cost Benefit Analysis manual(MLIT(2003)) . We suppose drive time is increase because of traffic jam and diversion
traffic by road construction. We estimate the difference of drive time with measureconstruction and without.
Drive cost reduction is also based on the same Cost Benefit Analysis manual.
The target contents is except concerned with drive time, of the cost reduction by drive
condition is improved. We estimated the cost, gas, grease, tire, maintenance, anddepreciation.
Comfort driving is well ride and well handling operation. We can quantitative
estimate the effect of improvement of road condition. The method is Willingness To
Pay (WTP). WTP is the amount people would be willing to pay for the value. We
estimated WTP by Contingent Valuation Method (CVM). We did a paper survey of
10,000 people lives in Tokyo in 2005, and estimated amount for good road condition.
Environment load reduction contents are air pollution, traffic noise, landscape,ecosystem, and energy. Executing our project, roadside environment is improved. We
estimated five contents. CO2, NOx, PM, and traffic tremble was estimated by past89

literature and existent data. Traffic noise was estimated by survey.
Traffic accident reduction is estimated three contents. One is human damage
cost about driver, people in the car with driver, and pedestrian. Second is object
damage cost about vehicle and construction that occurred damage by traffic accident.
Third is variation in damages cost by traffic concentration.
Project Prioritization
Project prioritization is decided by evaluation of the best combination of eachbridge project. Index of project priority is Net Present Value (NPV). NPV isinvestment index, private enterprise use of project feasibility research. We define thebest investment plan is not only obtain to minimum of project cost, but also achieve
NPV>0 and obtain to maximum of the difference between benefit and cost. We
developed loop program to decide project prioritization. This program can select thebest project combination to maximum of NPV for 30 years. The project combination is
calculated to prevent traffic concentration, by construction site with traffic restriction
is close to each other.
Asset Management System
We programmed application software which realized our plan, started
operation of the asset management system in 2007. The system is consisted of
calculation server, database server, web server, and firewall system. Server OS is
Windows 2000 server and database soft is oracle engine. The functions of the system
are database area and calculation area. By utilize accumulated data, the system cancalculate various programs, deterioration prediction, project prioritization, and so on.
Long-Term Bridge Management Plan
We made “Long-Term Bridge Management Plan” in 2009 with our asset
management methodology. We predicted deterioration of bridge and decided theadequate timing for measures in every bridge. Each measure, concrete deck exchange
to steel deck, steel girder to tied, and so on, set according to health index. The standard
costs of each measure refer latest rehabilitation technique. We evaluated Life Cycle
Cost (LCC) both corrective maintenance and preventive maintenance, and compared
two. Finally, we decided project prioritization and finished this planning.
This plan is all the bridges is divided into three maintenance types; long-life,
general and replacement. The measure to existent bridge, for earthquake and overloading, is also included.
The number of Long-life type maintenance bridge, is 212: high social value
bridge, and the bridge that rehabilitation is difficult for the construction is serious
affect to traffic. Kachidoki Bashi Bridge, Eitai bashi Bridge, and Kiyosu Bashi Bridge,
they are nominated national important heritage, is included. For the Long-life typebridge, measure to obtain the lifetime more than 100 years by utilized new technology
is executed.
Effects

90

Executing this plan produces five effects: cost, benefit, environment, safety,
and accountability.
First is the project peak cost cut and reduction of total project cost. The peakcost is cut by long-life measure is executed. The effect of the total project cost
reduction is expected 1,100 billion yen (1,200 million dollars, if one dollar is 90yen),
the total cost is 1,600 billion yen (1,800 million dollars) on corrective maintenance andthe total cost is 500 billion yen (600 million dollars) on preventive maintenance basedon this plan (figure 5).
Secondly, the best project combination plan can achieve to maximum of thedifference between benefit and cost. We try to minimum of construction cost anddevice the combination of construction project. We keep road safety and save thebenefit Tokyo citizens received.
The third is reduction of the load to environment. CO2 emission is reduced by
the life of bridge become long. Bridge replacement is decrease, quantity of usedmaterial becomes less, and the operation time of construction machine is decrease.
When the bridge management strategy is change from corrective maintenance topreventive maintenance, CO2 emission in thirty years is reduced 1,120 thousand ton,from 1,590 thousand ton to 470 thousand ton. For evaluating CO2 emission, we adopted
released unit as much as possible. We evaluated the emission data in each bridge, usedthe emission data of construction machine. The data is released from Japan Science
Committee Engineering and Center for Advanced Engineering Structual Assesment
and Research.
The fourth is save the road safety and secure. We can considerably loss the riskby the lack of safety. Because we estimate bridge condition by inspection, predict
deterioration and damage, and execute preventive measures. The safety of bridge issecured by the measures for long-life bridge. Because executing the measure forover-load and earthquake can improve durability of bridge.
The Final effect is progress of accountability. We execute the project whichbased on the plan utilized with the asset management system, we can account aboutplanning process and project assessment for Tokyo citizens.
Conclusion
This paper reported about the TMG measure for bridge management; try tochange management strategy from corrective maintenance to preventive maintenance.
“Long-term bridge management plan” have planned with developed Assetmanagement system. The purpose is bridge replacement peak cut, reduction of total
project cost, and save road safety. Some issues remain to steadily executing this plan:
secure the project organization system, reconfirm project cost, training in-houseengineers, and improving bridge inspection.
TMG forward preventive maintenance steadily, realizes effective and efficient
bridge management for Tokyo citizens, and keeps road safety and secure.
References
JSCE(2005), Challenge to Introduction of Asset Management, Gihodoshuppan, Tokyo,

91

pp.35-40 (in Japanese）
Ministry of Land, Infrastructure, Transportation, and Tourism (MLIT) (2003), Cost
Benefit Analysis manual (in Japanese)

92

270

架設橋梁数

Number of Construction

300

236

250

187

200

135

125

150

114
64

64

100
50

4

48

1
9 40

9 50

9 60

9 70

9 80

9 90

0 00

0 07

1- 1

1- 1

1- 1

1- 1

1- 1

1- 1

1- 2

1- 2

1 99

1 98

1 97

1 96

1 95

1 94

図1 東京都の橋梁建設年度別分布

2 00

9 30
1- 1

1 93

9 20

1 92

1- 1
1 91

9 10
1 90

1- 1

0

Construction
year (FY)
架設年度 (10年単位)
Figure 1: Distribution bridge construction year of TMG

RC
ＲＣ橋
8,494ｍ12％

Othersその他
12,054ｍ
16％

ＰＣPC
橋
12,256ｍ
17％

Othersその他
355,987㎡
32％

Steel
鋼橋
39,390ｍ
55％

橋
鋼Steel616,390㎡
56％

RC
ＲＣ橋
64,725㎡
PC
ＰＣ橋
6％68,419㎡
6％
橋
面積
Area

橋長
LengthOthers
その他
RC
ＲＣ橋
339橋
27％

98橋
8％

PC
ＰＣ橋
308橋
25％

Steel
鋼橋
502橋
40％

Number
橋数

Figure 2: classification by material of TMG bridge

93

昭和62年～平成17年（3次～6次） 橋梁の健全度割合の推移

0.7%

1987
昭和62年

37.7%

26.0%

27.7%

healthy
全
A ：健

8.0%

almosthealthy

B ：ほぼ健全

0.0%

1992
平成6年

36.3%

32.6%

21.4%

9.7%

littleattention

C ：やや注意
0.1%

1997
平成9年

26.5%

33.2%

30.3%

attention
意
D ：注

9.9%
0.4%

2002
平成14年

21.1%

0%

32.6%

20%

34.0%

40%

dangerous
険
E ：危

11.9%

60%

80%

100%

Figure 3: Progress of Bridge Health Index

5.0

REAL × YEARCURVE × YEAR

5.0
REAL × YEARCURVE × YEAR

Exponenti alType MetroCurve
4.5

4.5

Li n earTyp e M etroC urve

HealthRank = 5 exp(−0.00774 Li feTi me)
R2 =0.990

4.0

H ealthR ank = 5 − 0.27876 L i feTi m e

4.0

R2 =0.991
3.5

3.5

3.0

3.0

2.5

2.5

2.0

0

10

20

30

40

50

60

70

80

90

100

110

0

10

20

type 1:Exponential convex curve

40

50

60

70

5.0

CURVE × YEAR

4.5

2nd Polynomi alType MetroCurve

4.5

4.0

HealthRank = 5 −0.03235 Li feTi me+0.000047 Li feTi me 2

4.0

R2 =0.995

3.5

80

90

100

110

type 2: Linear type

5.0
REAL × YEAR30

REAL × YEARCURVE × YEAR

3.5

3.0

3.0

Mi yamotoType MetroCurve
2.5

HealthRan k = 5 − 0.00000237 Li feTi me 3

2.5

R2 =0.756

2.0

2.0
0

10

20

30

40

50

60

70

80

90

100

110

type 3: Second polynomial curve

0

10

20

30

40

50

60

70

80

90

type 4: Third concave curve

Figure 4: four alternative deterioration curve types

94

100

110

1,000

800
600
400
200
0
平成21
21

26
26

31
31

36
36

41
41

46
46

Reduction costfor 30years is
1,100 billion
50年度 yen

事業費（億円）

1,000
事業費（億円）

500 billion yen

1,600 billionyenLong-life
長寿命化対象橋梁

800

General and

一般管理対象橋梁・小橋梁
Replacement

600
400
200
0
21
平成21

26

26

31

31

36

36

41

41

Preventive
Maintenance

Corrective
Maintenance

Figure 5: Project cost compared preventive and corrective

95

46

46

50年度

blankINSTRUMENTATION AND MONITORING OF
I35W ST. ANTHONY FALLS BRIDGE
Catherine French1, Carol Shield1, Henryk Stolarski1, Brock Hedegaard2, Ben Jilk3
Abstract
The I35W St. Anthony Falls Bridge, constructed to replace the steel truss
bridge that collapsed in 2007, contains over 500 instruments to monitor the structural
behavior. Numerical models of the bridge are being developed and calibrated to thecollected data obtained during truck load testing and environmental loading. The dataobtained over the first few years of monitoring will be correlated with the calibrated
models and used to develop the baseline bridge behavior. This information will be
used to develop a system to monitor and interpret the long-term behavior of thebridge. This paper describes the instrumentation, preliminary results from the data andmodel calibration, and plan for developing the long-term monitoring capabilities.
Introduction
The I35W St. Anthony Falls Bridge, constructed to replace the steel truss
bridge that collapsed in 2007, consists of two parallel bridges to carry northbound andsouthbound traffic. The four-span bridges consist of prestressed concrete box girders.
Three of the spans were fabricated with cast-in-place concrete. The fourth span, theriver span (i.e., Span 2), was fabricated with match-cast precast segmental
construction. To accelerate the construction of the replacement bridge, the designbuild approach was used. The proposal from the design-build team of FlatironManson in conjunction with Figg Bridge Engineers featured the incorporation of a“smart-bridge” system. This system included instruments to monitor the structural
behavior of the bridge, as well as instruments to control the anti-icing and lighting
systems. Photographs of the bridge nearing completion of construction are shown in
Figure 1. Figure 2 shows elevation views of the bridge which indicate the primary
instrumented sections of the bridge and the variation in the cross-sectional shape ofthe boxes along the length of the bridge.
The University of Minnesota (UMN) is involved in the collection andinterpretation of the data obtained from the more than 500 sensors installed within thebridge as part of the “smart-bridge” system. In addition to using the information
obtained to better understand the behavior of prestressed concrete box girder bridges,
the UMN researchers are developing a system which can be used by the Minnesota
Department of Transportation to conduct long-term monitoring of the bridge.

1

Professor, Dept. of Civil Engineering, University of Minnesota, Mpls., MN
Graduate Research Asst., Dept. of Civil Engineering, University of Minnesota,
Mpls., MN
3Engineer, Minnesota Department of Transportation, MN
2

97Description of Instrumentation
This section contains a brief description of the types of sensors installed in thebridge followed by the quantities of each instrument in parentheses. Sensors include a
CorSenSys system (4), an example of which is shown in Figure 3, to monitor the
corrosion susceptibility of the reinforcement within the deck. Information from thissystem can be used to determine when the deck needs to be resurfaced to prevent thechlorides from reaching the reinforcement layer and initiating corrosion. Additional
instrumentation includes: strain gages to measure the deformations of the bridge (24resistive, 195 vibrating wire [VW], 12 fiber optic); accelerometers to measure thevibrations of the bridge and to relate those measurements to structural deformations if
possible (26); linear potentiometers to measure displacements of the expansion joints
in the structure (12), and thermistors to measure the thermal gradients through thebridge cross section (243, including the thermistors associated with the VW gages).
The vibrating wire strain gages and thermistor data are recorded statically at four totwenty-four times per day; whereas the accelerometer, resistive gage, and linear
potentiometer data are recorded at up to 200 Hz continuously during the day, and thenprocessed daily to save a shorter period of data coincident with the VW data, and theasynchronous major events of the day.
The strain, temperature, and vibration measuring systems in the bridge are
distributed in several sections throughout each span of the bridge. A few sections are
more heavily instrumented in order to investigate more detailed response which canbe compared to the limited set of information obtained in the other sections of thebridge. The exterior box of the southbound (SB) river span (Span 2) contains threedifferent types of systems (i.e., VW and fiber optic strain gages and accelerometers)that provide some redundancy and enable the evaluation of the relative effectivenessof the three different types of systems for consideration in future bridge monitoring
implementations. Examples of the more heavily instrumented sections are shown in
Figures 4 through 6.
Figure 4 shows the cross section of the SB bridge at midspan of Span 2 which
has the largest number of VW gages. The gages oriented in the longitudinal direction,
shown as solid circles in the figure, enable the measure of the longitudinal curvaturesat midspan and the distribution of the strains across the top flange of the bridge. Pairsof transverse gages, shown as solid rectangles, are located to investigate the transverse
curvatures at five locations across that section. The open circles in the figure indicate
locations of additional sets of thermistors, with the numbers in the circles indicating
the number of thermistors through the thickness at each location.
Figure 5 shows the SOFO (fiber optic) sensors which were located in Span 2of the SB exterior box. These twelve gages were oriented in pairs at six locations
distributed along the length of the span. Whereas VW strain gages have very shortgage lengths (i.e., ~6 in. [~150mm]), the SOFO gages measure strains over 13 ft [4m]gage lengths. The sets of SOFO gages are expected to provide information on theoverall curvature across the span which may be used to determine the deflections of
the SB river span.

98

Accelerometers are located below the deck near midspan of each of the boxes.
In the exterior box of SB Span 2, 14 accelerometers can be reconfigured in different
orientations at different locations. Figure 6 shows the typical location of theaccelerometers below the center of the deck (in all spans) and attached near the cornerof the flange in Span 2 (i.e., typical configuration of 13 of the 14 accelerometers inSpan 2). The accelerometers were attached near the corner of the flange to bettermeasure the overall dynamics of the span without the influence of the local deck
vibrations. The 13 accelerometers in Span 2 are currently distributed fairly uniformly
along the length, with one gage at midspan of the deck to replicate the configurationin the other spans.
Truck Load Tests
Prior to opening the bridge to traffic, a series of static and dynamic truck load
tests were conducted on the evenings of September 14 and 17, 2008. The loads were
provided by a series of eight heavily loaded sand trucks which were carefully weighed
and measured before the tests. The vehicles each weighed approximately the sameand had a combined weight of approximately 400 kips (1,800 kN). The trucks werepositioned in a series of different pre-established configurations on the bridge. For thestatic tests, the trucks were stationed at each of the pre-established locations longenough to typically capture three readings. Several of these tests were also repeatedover the course of the evening as time permitted. One of the configurations (ST I) is
shown in Figure 7, for which case the eight trucks were positioned across the width of
the bridge.
These tests provided valuable information used to calibrate numerical modelsof the bridge subjected to known loads at known locations. This information also
provided a “baseline” for the measured behavior of the bridge. If desired, future truck
tests could be conducted to compare the results to the initial baseline tests.
Finite Element Method (FEM) Model
Finite element models were developed to provide means to interpret the dataobtained from the bridge. The most current FEM model was created in ABAQUS
using continuum (solid) quadratic 20-node elements with reduced integration. The
benefits of the solid elements included the ability to accurately model the geometry of
the complex cross-sectional shape and to simulate the thermal gradient through thesection (node by node). The scope of the initial model was limited to the continuous
three-span section of the southbound bridge, as shown in Figure 8, where most of the
instrumentation was located. Boundary conditions were chosen to approximate thephysical constraints on the bridge including Piers 2 and 3 (assumed fixed at the baseand pinned at the top), with longitudinal expansion joints modeled at Abutment 1 andPier 4. To model the steel and prestress present in the bridge, all post-tensioning
tendons (with the exception of the draped external tendons) were approximated asshells embedded in the top and bottom flanges. Mild steel was assumed to beuniformly distributed throughout the section by adjusting the modulus of elasticity of
the concrete to account for the additional stiffness introduced by the reinforcement.

99All of the concrete in the FEM model was assumed to be normal-weight
concrete. The specified material properties for the cast-in-place and precast concrete
were initially used. Studies are currently underway at the University of Minnesota
using concrete samples obtained from the bridge during construction to investigate themeasured material properties of the bridge including modulus of elasticity, creep,
shrinkage, and coefficient of thermal expansion. The results of these studies may be
used to further refine the numerical models.
Calibration of FEM Model
The results of the truck load tests were used to calibrate the FEM model. The
results of the FEM model were compared to the measured data for the various
configurations and positions of the static truck tests. Figure 9 shows the longitudinalstrains and deformed configuration of the midspan cross section of SB Span 2(magnified 2500 times) with loading to simulate truck orientation ST I (as shown in
Fig. 7). Figure 10 shows the longitudinal mechanical strains obtained from the FEM
model at the top of the deck and 6 in (150mm) below the top of the deck, relative tothe VW strain gages assumed to be embedded at 6 in. (150mm) below the top of thedeck. From the figure, it is evident that the trends in the measured data are similar to
those obtained from the FEM model. The measured data better matched the FEM
model results predicted to occur slightly lower in the section (e.g., at 7 in. [180mm]
below the top of the deck). Differences between the measurements and the model may
be attributed to sources including differences in the as-built and assumed cross sectionof the structure (e.g., the deck thickness was estimated to vary by approximately 1in.),and potential errors associated with the as-built locations of the sensors embedded
within the structure.
In addition to calibrating the FEM model with the truck load tests, studies areunderway to calibrate the model with the measured results due to environmental
effects including the effect of the thermal gradient on the structure. Figure 11 showsthermal gradients measured through the cross section of the bridge obtained at fourtimes over an 18 hour period. The effect of solar radiation has a dramatic effect on thethermal gradient through the section particularly in the April to late June time frame.
Following calibration, the FEM model can be used to investigate potential
damage scenarios and how they might manifest themselves in the measured data.
Effects due to Temperature
To illustrate the effects of temperature on the response of the bridge, themeasured strains obtained at the top and bottom of the box section near midspan of
SB Span 2 are shown in Figure 12 over a twelve hour period during the course of oneof the truck tests. Note that the strain values in the plot were arbitrarily zeroed on
September 1, 2008; it is the changes in strain that can be determined from the plot thatare of importance. The peaks in the plot represent the strains when the load was
positioned locally with respect to the instrumentation. It should be emphasized thatthe truck tests took place over the course of an evening, from approximately 5:00 pm
to 5:00 am, so the effects of the solar radiation should have been minimized. From the100

figure, however, the temperature variation during that time frame can be noted by thechange in the readings of the bridge when it was unloaded. From this figure it isevident that the trucks caused a maximum local strain change of approximately 20εin the top gage readings.
Figure 13 shows the data from the same strain gages (i.e., in the top andbottom of the box section near midspan of SB Span 2) that were measured four times
a day (i.e., midnight, 6:00 am, noon and 6:00 pm) over a ten month period (arbitrarily
zeroed on September 1, 2008). As evident in the figure, the daily changes in totalstrain of the top gages (~100ε) due to the thermal gradient and temperature changes
were nearly five times the magnitudes of those obtained during the truck load testwith the eight trucks stationed across the bridge at this section (i.e., ~20ε). The
seasonal changes in total strain were more than 500ε for the same gages.
Besides having a significant effect on the strain measurements, thetemperature changes also appear to affect other bridge properties including the modal
frequencies. Preliminary data obtained from the accelerometer at midspan of Span 2
was used to determine the modal frequencies. To obtain this data, input averaging of
20 points was used on the dynamic data collected at 200 Hz over an approximately
half hour period to obtain greater resolution in the frequencies in the 0 to 5 Hz range
where the structural frequencies were expected to reside. The resulting Fast Fourier
Transform (FFT) applied to the data showed three strong peaks at approximately 0.8
Hz, 1.5 Hz, and 2.3 Hz as shown in Figure 14, which compared reasonably well withthe data obtained from the FEM model. In the FEM model, the frequencies around 0.8
Hz and 1.5 Hz represented bending modes, while the mode at 2.3 Hz was associatedwith a torsional mode. The FFTs were then applied to the data bimonthly, toinvestigate the consistency in the three noted frequencies over time. Figure 15 showsa sample of these results. The variation of the modal frequencies for the first mode isshown compared with the bridge temperature at the time of measurements. Dataobtained for the other modes showed similar variations with respect to temperature. It
appears that there is some correlation between temperature and bridge modal
frequencies. As the temperature increased, all modal frequencies were observed todecrease. Further studies regarding this behavior are currently underway.
Plan for Development of Long-Term Monitoring System
The data obtained over the first few years of monitoring the I35W St. Anthony
Falls Bridge will provide a “baseline” that describes what is considered “normal”
behavior of the bridge; however using the baseline data to establish absolute
maximum and minimum bounds on the expected behavior of the bridge is not
sufficient to identify abnormal behavior. As evident in the discussion of the measured
data above, the environmental effects (i.e., temperatures and thermal gradients) have asignificant impact on the bridge response. As noted in Figures 12 and 13 showing the
top longitudinal strain data measured near midspan of SB Span 2, the thermal effects
were observed to be approximately 25 times larger than the effect of eight fully
loaded sand trucks stationed across midspan of the SB bridge. In order to provide a
useful long-term bridge monitoring system, it is important to be able to discern the101relationship between the thermal effects and the behavior of the bridge. The desire isto provide a range of “moving” bounds that are related to the range of expected results
associated with the measured thermal gradients. The FEM model which has beencalibrated to both the truck load data and the data associated with thermal effects willserve as a useful tool in this regard. Any changes in expected behavior outside of the“moving” bounds will provide a means to signal when the response of the bridge mayneed further investigation or when maintenance needs to be performed.
One of the challenges is that no “turn-key” system for monitoring bridges suchas this exists. The monitoring system has to be created where the inputs from thebridge (i.e., measured temperature distributions and expected traffic loads) can be
combined with other information (e.g., measured strains) to identify anomalies inbehavior. In order to provide a range of “moving” bounds, the calibrated FEM model
will be used to establish relationships between the thermal gradient effects and the
expected responses from the sensor data. As an example, a “look-up” table may be
developed that could correlate the measured data obtained from the thermistors to themeasured data from the other types of sensors (e.g., strain gages and linear
potentiometers).
Figure 16 shows a schematic that illustrates the plan for the structuralmonitoring system under development. In the schematic, the raw data from themultiple dynamic and static acquisition systems is collected and analyzed orprocessed. The outputs of the processed data are then considered either model input(e.g., thermal effects) or response (i.e., measured responses of the bridge associatedwith the measured model inputs). The model input is used to determine expected
response (e.g., expected bridge curvatures, deflections, etc.), by means of a “passive”
or “active model.” A “look-up” table, as described above, is an example of a“passive” model. An example of an “active” model would be a study with thecalibrated FEM model to investigate an intermittent load test of the bridge. The
expected responses from the active or passive models would be compared to themeasured “response” of the bridge. The system would then output the results of thecomparison to the bridge management engineer. Example outputs include
notifications when the measured responses are out of range of the expected responses
which may warrant further examination.
The system will also be designed to notify the bridge management engineer
when any problems are encountered with the data collection system, such that steps
may be readily taken to avoid the loss of information. One of the challenges with thelarge volume of data collected with the system is the large amount of storage required
to retain data. The system under development will be designed to cull the collected
data to maintain sufficient information to ensure gradual changes in behavior are
documented without causing overwhelming long-term storage demands.

102

Summary
A long-term structural monitoring system is being developed by the UMN for
the I35W St. Anthony Falls Bridge. This structure contains over 500 instruments
which includes sensors to monitor corrosion susceptibility, accelerations, strains, andmovements at the piers and abutments. Numerical models of the bridge have beendeveloped and calibrated to the collected data obtained from truck load tests, while
studies are currently underway to calibrate the model to environmental (thermal)
loading. The data obtained over the first few years of monitoring will be correlatedwith the calibrated models and used to develop the baseline bridge behavior. Because
the thermal effects have such a significant impact on the response of the bridge, oneof the challenges in developing the long-term structural monitoring system is thecreation of “moving bounds” to distinguish when the response of the bridge is out of
the expected range. The long-term monitoring system is being designed as a tool for
bridge management engineers to investigate when the response of the bridge requires
further evaluation and when maintenance may need to be performed.
Acknowledgments
This research is sponsored by the Minnesota Department of Transportation(Mn/DOT). The views expressed herein are those of the authors and do not necessarilyreflect those of the sponsors.

Unit Conversion
1 in. = 25.4 mm
1 k = 4.448 kN

FIGURE 1: PHOTOS OF THE I35W ST. ANTHONY FALLS BRIDGE DURING
CONSTRUCTION

103

1225ft (373m)

500ft (152m)

(a)

(b)
FIGURE 2: ELEVATION VIEWS INDICATING (a) INSTRUMENTED
LOCATIONS AND (b) CROSS-SECTIONAL SHAPES ALONG THE BRIDGEFIGURE 3: EXAMPLE OF CORSENSYS CORROSION SENSOR
INSTALLATION

104

FIGURE 4: VIBRATING WIRE STRAIN GAGE LAYOUT FOR MIDSPAN OF
SPAN 2 OF SOUTHBOUND BRIDGE (LOCATION 7SB)

FIGURE 5: SOFO SENSOR LOCATIONS ALONG LENGTH OF SPAN 2 OF
SOUTHBOUND BRIDGE EXTERIOR BOX

FIGURE 6: ACCELEROMETER LAYOUT IN THE SOUTHBOUND BRIDGE

105

FIGURE 7: TRUCK ORIENTATION ST I

FIGURE 8: FINITE ELEMENT MODEL OF I35W ST. ANTHONY FALLSBRIDGE106

FIGURE 9: LONGITUDINAL STRAINS AT MIDSPAN OF SPAN 2 UNDERTRUCK TEST STI-7SB (DEFORMATIONS MAGNIFIED BY 2500)

FIGURE 10: LONGITUDINALMECHANICAL STRAINS ACROSS SECTION(MIDSPAN OF SB SPAN 2)

107

FIGURE 11: MEASURED THERMAL GRADIENTS THROUGH THE SECTION
DEPTH OF WEB (MIDSPAN OF SB SPAN 2) 4/9/2009

FIGURE 12: MEASURED TOP AND BOTTOM TOTAL STRAINS (MIDSPAN OF
SB SPAN 2) AND TEMPERATURE VARIATION VS. TIME DURING TRUCK
TESTS

108

FIGURE 13: MEASURED TOP AND BOTTOM TOTAL STRAINS (MIDSPAN OF
SB SPAN 2) VS. TIME (OVER 12 MONTH PERIOD)

Frequency Spectrum: 11/1/2008 12:41:29 PM
Node 6, Channel 1: Midspan of Span 2, Exterior
Box
3.00E‐05

Magnitude

2.50E‐05
2.00E‐05
1.50E‐05
1.00E‐05
5.00E‐06
0.00E+00
0

0.5

1

1.5

2

2.5

3

3.5

4

Frequency (Hz)FIGURE 14: MODAL FREQUENCIES OF SPAN 2 ON 5/9/2009

109

4.5

5

FIGURE 15: MODAL FREQUENCIES OF SPAN 2 (OVER EIGHT MONTHPERIOD)

FIGURE 16: SCHEMATIC FOR DEVELOPMENT OF STRUCTURALMONITORING SYSTEM

110

Study for Bridge Renewal and Repair by Osaka Municipal Government
Shinsuke Yumoto1
Tetsuya Yokota2
Yasutomo Komatsu3
Abstract
There are 764 bridges which managed by Osaka Municipal Government (incl.
about 100 old bridges that over 70 years old)
Although many old bridges will be 100 years old in coming 30 years, to replace
the all old bridges within 30 years is impossible when consider the financial pressure in
Osaka city. Therefore, it provided the replacement plan refereed to the Study Group.
Specifically, Osaka Municipal Government created the own Replacement
Judgment Policy through the Evaluation of Health Assessment by bridge inspection
and Evaluation of Functionality by structure collation. This is the report of the
Judgment Process.
Introduction
Osaka is the second big city next to Tokyo
where has an area of about 222 km2, population isabout 2,600,000 that is the metropolis in the westernpart of Japan (Fig.1).
There are many bridges which loved bycitizens such as Tenjinbashi-Bridge,
Tenmabashi-Bridge and Naniwa-Bridge (Naniwa
major three bridges). Osaka is called “Naniwa
Happyaku-Ya-Bashi” whereas Tokyo is called “EdoHappyaku-Ya-Cho”.

RussiaChina

North Korea
Korea
Tokyo

Osaka
Osaka Municipal Government is managing
about 764 bridges (Bridges Area is about
720,000m2 (2009.1st. Apr.), including Consecutive
Fig.1 Location of Osaka
viaducts (such as Midosuji to have traffic density
more than 100,000 per day) (Photo. 1), long bridges over the vast river, well known
bridges (Photo. 2), and many kinds of bridges.
On the other hands, many bridges were constructed during the First City
Planning Stage, started from 1921, the ratio of old bridges which over 50 years old
1

Public Works Bureau, Osaka Municipal Government, Osaka, Japan
Public Works Bureau, Osaka Municipal Government, Osaka, Japan
3
Public Works Bureau, Osaka Municipal Government, Osaka, Japan
2

111

holds about 20% of all bridges. Compared with about 6% of the Japanese average, it
indicates that how the aging of the bridge progress in Osaka city. Furthermore, about
100 bridges will become 100 years old in coming 30 years (Fig.2). To provide of themaintenance principle for these old bridges becomes the urgent important problem.
180
160

Osaka Municipal Government is
considering maintenance and thereplacement plan with the Study Groupon the Bridge Maintenance and Renewalof the Osaka Municipal Government
from 2007. This is the report of themaintenance process of the old bridges
through this examination.

Osaka Expo（about 1970）

Number of bridges

140120
about 100 bridges will become
100 years old in coming 30
years

100
80
60
40
20

the First City Planning
Stage（About 1930）

1-5

6-10

11-15

16-20

21-25

26-30

31-35

36-40

41-45

46-50

51-55

56-60

61-65

66-70

71-75

76-80

81-85

86-90

91-95

0

Age of bridges

Fig.2 Distribution of the Number of Bridges(As of Apr.1, 2008)

Photo 1 Shinmido Viaducts

Photo 2 Naniwa Bridge

Recently, Osaka Municipal
Government is very severe financial
status. Although it controlled to build
the new bridge and replacement suchas reconstruction costs or theearthquake proof construction, cost
reduction by new maintenancetechnique is demanded. It becausethat the conventional expense
measure was insufficient. (Fig.3)

Project Cost
事業費(億円）（hundredmillion yen）

The background of the examination
橋梁事業費の推移

100
90
80
70
60
50
40
30
20
10
0

Reconstruction改築費(耐震除く）
costs （exc. Aseismic ）改築費(耐震除く）
Aseismic costs 耐震費
耐震費
Maintenance costs
維持費
維持費

S63

H1

H2

H3

H4

H5

H6
H6

H7
H7

H8
H8

H9
H9

H10
H10 H11H11H12H12 H13H13 H14
H14

H15
H15

H16
H16 H17H17

H18 H19
H19
H18

年 Year度※Inc.
New, Reconstruct and Provision of the large-sized vehicle in Reconstruction costs （exc. Aseismic ）※改築費(耐震除く)には、新設、架替、車両大型化対策が含まれる

Fig 3 The Process of the Bridge Business Expense

Osaka Municipal Office has started to operate the Bridge Maintenance System(OBMS) since 2005. OBMS was developed from 2003 which introduced technique ofthe asset management.
Although Osaka Municipal Office is basis on the bridge prolongation by the112

prevention maintenance, it found out that replacing the minimum percentage of bridge
is very important. Therefore, it required classify bridges into prolongation andreplacement to consider the maintenance and the replacement plan for old bridges.
Next contents are explaining the process of Replacement judgment.
Replacement judgment policy of old bridges
As an object of 100 super old
bridges older than 70 years old whichconstructed before WWRII (except thebridges which have replacement or
removal plan), Osaka city had made the
1st selection for bridges which
necessary the detail analysis throughthe Replacement judgment matrix.
Then, evaluated general evaluation
based on LCC analysis and selected
bridges which are necessity renewal ornot (Fig.4).

Fig.4 Diagram of Replacement Plan

Replacement judgment matrix(1) Evaluation method
It is common way to consider the maintenance plan based on the result of by thebridge inspection when made the plan for maintenance plan.
However, the judgment on the bridge replacement only on the basis of thebridge health assessment may lead to wrong decisions. This is caused by the view ofthe fact that the old bridges built before the World War II were designed using different
codes from now with respect to loads, earthquake-resistant design and river conditions
and do not correspond to the present design codes.
For the reason, the final decision of replacement of old bridge was judged to use
evaluation both 1.Evaluation of Health assessment and 2.Evaluation of Functionality.
(2) Evaluation of Health assessment (by bridge inspection)
As for the evaluation of healthassessment, Osaka city evaluated it on
the basis of future deterioration progress
based on particularity inspection(nearness viewing) and the result of
Nondestructive testing (Fig.5).

Particularity Inspection

Nondestructive Testing

Evaluation of bridge Inspection
I…No Damage
II…Slight Damage
III…Damage is not small
IV…Big Damage** Problem bridges based on the strength of concreteand carbonation in the slabs, bridge piers and abutments.

Fig.5 Evaluation of Health Assessment

About the Nondestructive testing, it is the quantitative evaluate which cannot to113

confirm by viewing kind of the measurement of the neutralization depth and the saltcontent and concrete strength testing. It could be anticipate the precision improvementof the evaluation of health assessment.
Furthermore, I-III was judged to use status index1) by National Institute for Landand Infrastructure Management as trial. Evaluation IV that the damage was the mostremarkable defined it that concrete strength deteriorated remarkably.
(3) Evaluation of Functionality (by Structure collation)
The functional evaluation was
instituted whether load resistance, seismic
resistance and river conditions could takeadjustment for a current standard (Fig.6).
The river conditions are impediment ratioof river flow, standard diameter lengthand the height under the girder.

Load carrying capacity

Aseismic
Consistency ofriver conditions

Evaluation of structuralfunctionality
I…No Deficiency
II…Partial Deficiency
III…Deficiency is not small
IV…Big Deficiency** Raise (in case of steel pile-support); Masonry Abutment; Liquefaction;
Extremely small clearance; Bad Road Alignment

Fig.6 Evaluation of Functionality

About the bridge which road linear shapes extremely bad were evaluated IV as
the bridge that lack of function was remarkable such as pilling-stones abutment(Photo.3), the liquefaction ground, the bridge which raises a figure (Photo 4) and thebridge that a girder soaks in water at the time of rise of the river.

Raising bridge floor bottom

Photo3 E.g. of Masonry Abutment

Photo4 Raising Bridge Floor Bottom by a
Steel Beam.

(4) Extraction result of bridge renewal examination

Lack Level of Function

According to the result of thereplacement judgment matrix, thebridges which did not satisfy a current

Evaluation of Functionality

The bridges divided into to two groups “The bridges for prolongation” and “Thebridges to be replaced” through theHighIVresult of both evaluation of Healthassessment and Functionality. Fig.7 is
The bridges to be replacedIIIshown that the Replacement judgment
matrix.2)
II

The bridges for prolongation
Low I
I
Small

II

III
Level of Damage

Evaluation of health assessment

IV
Big

Fig.7 Replacement Judgment Matrix

114

standard were chosen as the bridges renewal examination which should have examinedreconstructs in future despite of the soundness was no problems in inspection. The
standard is changing with the times, so that old bridge does not adapt the current
standard properly. On the other hand, many of old bridges which kept soundness
enough even now were confirmed.
From these results, there is a limit to judge the condition of the bridge from onlya check result by the viewing. It means that admitted it is necessary to examine acharacteristic each bridges in detail when makes an important decision for replace andso on.

General evaluation based on LCC Analysis
The economical evaluation wasassessed by LCC of the next 50 years
which based on the result of replacement
and prolongation analysis in thereplacement detail judgments （2ndselection) of super old bridges.(Fig.8) In
addition, it were judged whether necessity
the replacement or not which basis of theconsistency of higher plan. The next
expression is shown the detail of LCC.

Fig.8 Idea of Life Cycle

LCC = [Replacement Costs] or [Reinforcement (Improvement) Costs] +[Maintenance & Management Costs] + [Traffic congestion losses due to construction]+ [Risks] - [Benefits] - [Survival Values]
While Replacement costs and Maintenance & Management costs are evaluation
objects normally, Risks, Benefits and Survival values were appropriated as costs in thisexamination, then, the replacement bridges are judged objective and quantitatively.
[Replacement costs] was calculated based on structural calculation and
Maintenance & Management costs. In the case of reconstruct, as those old bridges arestill usable even after 50 years, [Survival values] is to be considered. Additionally,
traffic regulation is required when renewal the bridges, therefore, [Traffic congestion
losses due to construction] need to be considered. Traffic congestion loss wascalculated by carrying out traffic estimation using traffic network assignment of all areaof Osaka City, then calculated 3 benefits such as traffic time, traffic costs and loss of
traffic accident basis on the current situation and discrepancy of traffic regulation.
As for [Reinforcement & Improvement costs], it calculated to consider theseismic reinforcements, the necessity of the anti-vehicle upsizing measure andreinforcement structures. In the case of the shiftlessness, it assumed that appropriatedthe amount of damage when an earthquake occurred as [Risk]. It is attributed to be115

damaged at the time of an earthquake.
[Benefits] was appropriated it when the function of the bridge improved in bothcase of renewal and reinforcements and improvements.
The case of Test Calculation

Prolongation
Plan
Replacement
Plan
Shiftlessness
plan

LCC Ratio

A bridge is the Gerber girder
bridge which has the wooden postsof about 80 years. Although thegirder had rose, it required thefoundation reinforcement in the
result of the seismic verification.
However, according to the LCC
analysis, the reconstruct plan is
economical rather than foundation
reinforcement. In shiftlessness plan,
the earthquake damage were
occurred sort of traffic suspension
or the recovery costs, besides theamount of damage by stops of the
production activity will increase.

Present
Condition

3.0
2.5
2.0
1.5
1.0
0.5
0.0

A Bridge
Gerber girder bridge,
Wooden posts
Wooden Postsin
Liquefaction Ground,Less seismic
Pier,
Basic reinforcement
Steel Flooring Bridge

Damagefrom Earthquakes

2.5

B bridge
Arch Bridge

No rooms under thegirder, obstruction of thewater transportation
Floor reinforcement

Arch Bridge
Only repair

Amount of earthquake damage
Traffic congestion losses due to construction
Maintenance & Management Costs
Reinforcement (Improvement) Costs
Replacement Costs

without

LCC Ratio

Fig.9 is shown the examples
that the result of LCC analysis of
the bridges which chosen
prolongation or reconstructed. As
comparison, in case of minimum
maintenance and management of
LCC was shown without
reconstruct and prolongation
measure (the following"Shiftlessness plan"). The mostsuitable plan of LCC ratio is 1.0.
The cost of reconstruct is subtracted
for survival value.

Prolongnation Plan

Replacement

Amount of water transportation loss
Amount of earthquake damage
Traffic congestion losses due to construction
Maintenance & Management Costs
Reinforcement (Improvement) Costs
Replacement Costs

2.0
1.5
1.0
0.5
0.0
without

Prolongnation

Replacement

B bridge is the arch bridge,
Planseismic is secured, used about 80
Fig.9 Result of LCC Analysis
years. While the project of thewater transportation activation is promoted by the Water Metropolis Revival Plan,
there is few girder bottom room of the bridge it becomes the obstruction of the watertransportation, so that the water transportation loss was appropriated. This is caused by
the girder bottom room was not secured by the refinement construction work orreconstruct plan in Shiftlessness plan.

116

As a result of having evaluated LCC or consistency with the higher plan in
object bridges, it is judged the prolongation plan is the most suitable plan through therefinement construction work of the upper side of the bridge. Moreover, reconstructplan were assumed as the same type of now, because of the well known bridges havehistorical values.
Conclusion
Osaka city is considering themaintenance and the replacement plan of
the old Bridges based on a basic policy.
Fig.10 is shows the result of these
examinations

Old Bridges (about 100）

Planning Replacement / Removal or
not by other plans

1st Selection（Replacement Judgment Matrix）(1) As the result of the 1st
selection by Replacement judgment
matrix, about 20 bridges were selected as
an object of replacement judgment fromabout 100 old bridges except bridges
which were to be rebuilding by otherplans.
(2) As the result of the 2nd
selection by LCC analysis, about 10
bridges were selected after the 1st
selection.

Past
Examinations

2nd Selection（LCC・General Evaluation）

The precision improvement of the
Plan（The inspection of the local
condition, the enforcement of the

FutureExaminations

Fig.10 Schedule of Examinations(3) To other object of reconstructbridges is to be trying for the precision
improvement of the project plan basedon the enforcement of the close
inspection of the local condition and thedetailed design in future including
valuable in the history of civil
technology (Photo.5,6) and so on. Osaka
city will also examine how manage thesebridges in the future.

Photo 5 Suisho Bridge(Well-known Bridge)

As regards to the plan theory of
LCC analysis when consider thereinforcements or replacements, it
investigated with various way3)4)5). This
examination was important to provide
the maintenance plan of bridges on thebasis of the results of the past studies.

Photo 6 Honmachi Bridge(Oldest Bridge built in 1913)

117

Acknowledgments
This is the report that explaining the basic concepts regarding the project on thereconstruct of old and historical bridges. Thanks to the well-preserved drawings,
documents of summary of works and records of works on the bridges built before theWorld War II, the project has been promoted quite smoothly. Thus, the effort andtradition of the bridge engineers of the Osaka Municipal Office must be highly
respected. The importance of the succession of the important documents andinformation to the next generations is becoming more and more important.
Finally, we would like to take this opportunity to express their appreciation to
the member of the Study Group on the Bridge Maintenance and Renewal of the Osaka
Municipal Government for their precious opinions and advice., Chairperson: Mr.
E.Watanabe (Emeritus of Kyoto University), Committee: Mr. H. Furuta(Professor of
Kansai University) and Ms. M.Tanaka (Assistant Professor of Osaka Industrial
University).
References
1) Hiratsuka, Tamakoshi, Takeda, Tateyama, Study for general evaluationindex of Road Bridges (61st Annual Meeting of JSCE 2006.9)
2) Yokota, Komatsu, Nagahashi, Replacement Plan of Old Bridges by Osaka
Municipal Government (Bridges and Foundation, 2008.8)
3) Kaito, Abe, Kumon, Fujino, Bridge management based on minimization of
total cost considering its stock value (Proceedings of Structural Engineering 2001.3)
4) Yamaguchi, Ito, Miki, Ichikawa, Life Cycle Cost evaluation of highway
bridges considering social influence (Proceedings of Structural Engineering 2001.3)
5) Sato, Yoshida, Masumoto, Kanaji, Prioritization of seismic retrofit for roadbridges based on the concept of Life Cycle Cost (Proceedings of JSCE 2005.3)

118

25th US-Japan
Bridge Engineering Workshop

Session3

Accelerated Bridge Construction
Seismic Performance and Structural Details of Precast Segmental Concrete Bridge Columns
By Jun-ichi Hoshikuma, Shigeki Unjoh, and Junichi Sakai
Accelerated Constructions of the Viaducts on the Second Keihan Expressway
By Yoshihiko Taira, Hirotsugu Mizuno,Kei Muroda, and Akio Kasuga
New FHWA Seismic Hazard Mitigation Studies for Highway Bridges
By W. Phillip Yen
Overview of the Development Process and Effects of the UFO Method
By Yuji Mishima, Yukio Katsuta, and Tomoaki Tsuji

Page Left Blank

SEISMIC PERFORMNACE AND STRUCTURAL DETAILS OF PRECASTSEGMENTAL CONCRETE BRIDGE COLUMNS
Jun-ichi Hoshikuma1, Shigeki Unjoh2, Junichi Sakai3

Abstract
The precast segmental concrete bridge column would be one of the options forthe accelerated bridge constructions because the construction period at site can beshorten due to no need of formwork, placement and curing of concrete. Thus, thosecolumns are expected to be applied for the bridges for overpass crossings in urban
areas to minimize the effect on existing traffic. Additionally, high quality of theconcrete members would be ensured because the concrete segments are fabricated atfactories. This paper briefly introduces the state-of-practice of the segmental concrete
bridge columns in Japan. Furthermore, recent research activities in PWRI for the
precast segmental concrete bridge columns are summarized.
Introduction
With a background of the generalization of the performance-based designconcept into practices, the applications of new materials, new designs, and new
structures have initiated to be actively employed with necessary performanceverifications. The precast segmental bridge columns are one of such new applications,and effectively use the combination of high strength materials including steels andconcrete. The precast segments are fabricated at factory, so that the precast segmental
bridge columns can be easily achieved to have better-quality. Therefore, the precast
segmental bridge columns are expected to improve the constructionability at sites andshorten the construction period.
Public Works Research Institute conducted 2-years joint research program forthe development of the new precast segmental concrete bridge columns with threeprivate construction companies. In the research program, three types of structuraldetails of precast segmental concrete bridge columns were proposed and the failure
mechanism of the proposed columns was investigated through a series of shake table
test. Based on the experimental studies, the limit states for the required seismic
performance are discussed in this paper. Structural details for those columns were also
introduced.

1

Chief Researcher, Center for Advanced Engineering Structural Assessment andResearch, Public Works Research Institute (PWRI)
2
Research Coordinator for Earthquake Disaster Prevention, Research Center for
Disaster Risk Management, National Institute for Land and InfrastructureManagement (NILIM)
3
Senior Researcher, Center for Advanced Engineering Structural Assessment andResearch, Public Works Research Institute (PWRI)

121

Structural Concept of Precast Segmental Bridge Columns
Figure 1 shows the outline of the precast segmental bridge columns which havebeen designed and constructed in the past in Japan. The precast segments are produced
at factory and transported to the construction site. These segments are piled up at thesite and connected each other through the steel bars, to be a column. It would be an
important advantage to shorten the construction period at the site because of no needof formwork, placement and curing of concrete. Therefore, the precast segmental
bridge system would be expected to be applied for overpass crossings in the urban
areas in order to decrease the traffic jamming and then to minimize the effect on
existing traffic.
Structural details of the conventional precast segmental bridge columns areshown in Figure 2. There are two types of the precast segmental columns. The precast
PC columns are built with the segments connected through the high strength steel bars
columns, as illustrated on the left side section in Figure 2. Each segment ispost-tensioned by all high strength steel bars, to integrate with column structure. After
post-tensioning, the following segment is piled up on the lower segment and the highstrength steel bars are installed into the section through the ducts. These bars arecoupled with the lower high strength steel bars and the grout is injected to the duct tobe bonded. Details of the segment connection are shown in Figure 3. These processes
are repeated up to the column height.
On the other hand, the precast RC
columns are built with the segments
connected basically through the nominal
strength steel bar, as illustrated on the rightside section in Figure 2. A few high-strength
steel bars are installed and minimum
post-tension required for setup of the segment
is applied to these bars. The steel bars arecoupled and grouted with the same procedureas the precast PC column.
Construction samples of the precast
segmental concrete columns or piles areshown in Photos 1 and 2. Photo 1 showsconnection of the precast segmental PC piles
with the cast-in-place footing. Photos 2
shows the precast RC oval column
constructed very close to existing bridge
column.
Figure 1 Illustration of Precast Segmental
Concrete Columns

122Figure 2 Concepts of Connection Details in Conventional
Precast Segmental Concrete Columns

Figure 3 Connection Details of Steel Bars in Segment

123Photo 1 Connection of Precast Segmental PC Piles with Cast-in-place Footing

Photo 2 Precast RC Oval Column Constructed Very Close to Existing Column

Recent Research Activities for Precast Segmental Bridge Columns in PWRIPWRI conducted the joint research program on the new precast segmental
concrete columns with 3 private companies including Kajima Co., Sumitomo Mitsui
Construction Co., Ltd., and P.S. Mitsubishi Construction Co., Ltd. Three types of the
precast segmental concrete column details were proposed. Research issues were to124

obtain the data on the failure mechanism, the strength and ductility performance, andthe dynamic behavior of proposed precast segmental bridge columns, and to develop
the design method including the limit states to achieve necessary seismic performance,
detailed design methods for segments, joints, PC cables, bending–shear resistanceevaluation, and construction methods. In the joint research program, a series of cyclicloading tests, shaking table tests, and analytical studies were made to develop theseismic design guidelines for the proposed precast segmental concrete columns.
Structural concepts of the proposed precast segmental concrete columns areshown in Figures 4 to 6. The structural details and properties are described in thefollowings. Figure 4 shows the outline of the precast PC column proposed by Kajima
Co. The segments are piled up at the site, and each segment has the outer and inner steel
pipes. Outer steel piles are embedded in the segment when it is produced at factory, andinner steel pipe are installed at the site between the segments. The inner steel pipe is toresist against the shear force acted the joints of segments as a shear key. After the pilingup of all segments for columns, the vertical tensioning force is applied for segments bypost-tensioning bars through the inner pipes, in which the those bars work as
longitudinal steel.
Sumitomo Mitsui Construction Co., Ltd proposed the precast segmental PC
hybrid columns as shown in Figure 5. The segments are piled up at the site. Eachsegment is made of combination of inside steel shell and outside concrete. Inside steel
shells of the segments are connected by steel bolts. After the piling up of all segments
for column, vertical tensioning is applied for segments by external post-tensioning bars.
At the joints between segments, shear keys are provided at the edge of steel shell andconcrete mortar is placed between the segments outside concrete. Therefore, verticalaxial force by dead load and live load is carried by inside steel shell but the earthquake
force is carried by steel shells, bolts, and outside concrete. Shear force acted at thejoints is carried by the shear keys of steel shell. Steel shells and bolts, and PC cables
works as longitudinal steel. The bolts are designed to be firstly yielded when the
deformation is exceeding the elastic limit and then the steel shells are not expected to
be damaged. It is an important concept for this column that the yielded bolts can be
replaceable after the earthquake and then the columns can be easily recovered to theoriginal performance.
Figure 6 illustrates the precast RC column proposed by P.S. Mitsubishi
Construction Co., Ltd. The concrete segments are piled up at the site with a few
temporally post-tensioning bars. Those bars are provided not for tensioning as
longitudinal steel but for just construction to assure the quality of the joint connection
between segments by resin. After piling up of all segments, mild longitudinal steel bars
are inserted into the ducts, which are pre-grouted with the mortar from the top to thebottom of the column. The columns are made of segments but the design concept is thesame as nominal reinforced concrete column. Since the longitudinal steel bars areplaced inside the sheathe of segments, so the confinement effect to prevent thebuckling of longitudinal bars is much higher than the nominal reinforced concrete
columns which is confined by the cover concrete and lateral steel bars.

125Figure 4 Precast Segmental Prestressed Columns Proposed by Kajima Co., Ltd.

Figure 5 Precast Segmental Hybrid Columns Proposed by Sumitomo Mitsui Co., Ltd.

Figure 6 Precast Segmental RC Columns Proposed by P.S. Mitsubishi Co., Ltd.

126

Limit States of Precast Segmental Bridge Columns
General Concepts of Limit States
It should be essential to determine the limit state for each seismic performancelevel, to develop the seismic design method. In Design Specifications for Highway
Bridges issued by Japan Road Association, three seismic performance levels are
specified for intensities of design ground motions and importance of the bridges. The
limit states are determined based on the required seismic performance, which aredescribed in terms of the safety, serviceability and repairability. The schematic imageof the limit state for the conventional reinforced concrete bridge column is shown in
Figure 7.
For the seismic performance level 1, it is required to ensure the normal
functions of bridges after an earthquake, which means the mechanical properties of thestructural members should behave within the elastic ranges. For each structuralmember, the stress induced by an earthquake shall not exceed its allowable stress. For
the seismic performance level 2, it is required to ensure the serviceability andrepairability after an earthquake. As the limit state, the structural members in which thenonlinear behavior is allowed deform beyond elastic range but within a range of easyfunctional recovery. For the seismic performance level 3, it is required to ensure thestructural safety during an earthquake. Since neither serviceability nor repairability isrequired, structural members in which the nonlinear behavior is allowed deform within
the ultimate ductility capacity.
Limit States for Proposed Precast Segmental Concrete Bridge Columns

Lateral force

The limit states for the precast segmental concrete bridge columns shown in
Figures 4, 5 and 6 are determined considering the structural properties and nonlinear
behavior of each structure. For the seismic performance level 1, the limit states of the
precast segmental structures are determined to be same to a conventional reinforced
concrete column. For each structural member, the stress induced by an earthquake shallnot exceed its allowable stress.
The limit states for the seismic performance levels 2 and 3 are determined based on the
nonlinear behavior of each structure, which are introduced in details below.

x: Before Yielding of Rebar: Minor Spalling of Cover Co.
x : Before Rebar Buckling
Lateraldisplacement

Figure 7 Limit states for Conventional Reinforced Concrete Columns

127

Limit States for Precast Prestressed Concrete Bridge Column
Because no mild longitudinal reinforcement is provided in the precast
prestressed concrete bridge column, the yielding of the post-tensioning steel bar is theimportant limit state. Once the yielding of the post-tensioning steel bar occur, it isdifficult to recover the required functions. Based on these properties of the structure,
the limit states shall be determined as shown in Figure 8.

Lateral force

For the seismic performance level 2, the yielding of the post-tensioning steel
bar is determined as the limit state to ensure the serviceability and repairability. Figure
9 shows the force-displacement hysteresis and the failure mode obtained from the
shake table tests for the precast prestressed columns during the design level earthquake
ground motion. The post-tensioning steel bars remained in the elastic range and minor
spalling of cover concrete was observed. Although the column model performed wellbeyond the yielding of the post-tensioning steel bars in the shake table test, the range
beyond this point is not considered in the seismic design of the precast prestressed
concrete bridge for safety consideration. Thus, the limit state for the seismic
performance levels 3 should be determined beyond the yielding of the post-tensioning
steel bars. Further research may be needed for consideration of the behavior after
yielding of the post-tensioning steel bars.

x: Elastic rangex : Before Yielding ofPrestressing Steel

Lateraldisplacement

Figure 8 Limit states of precast prestressed concrete columns
Lateral force (kN)
600
400
200
0
-200
-400
-600
-120

-80

-40

0

40

80

120

Lateraldisplacement (mm)
慣性力作用位置の変位（mm）

Figure 9 Seismic Response and Failure Mode after Design Ground Motion Level Test
for Precast Prestressed Column

128Lateral force

Limit States of Precast Hybrid Bridge Column
The schematic image of the limit states of the precast hybrid bridge column isshown in Figure 10. Since the key feature of this column is the repairability byreplacement of the connecting bolts, the limit states of the seismic performance level3 should be determined to be same as those of the seismic performance level 2. The
replaceable limit of the connecting bolts is determined as the limit state for the seismic
performance level 2. The allowable strain of the bolts is estimate to be 2% based on the
low-cycle fatigue tests and the shake table tests. The other structural members should
remain in the elastic range. Figure 11 shows the force-displacement hysteresis and thefailure mode obtained from the shake table tests during the design level earthquake
ground motion. The results from as-built series (case 1) and post-repair series (case 2)
are compared. After the design level tests in the post-repair series, the response
displacement was still smaller than the design displacement while about 2% strain was
induced in the connecting bolts and minor spalling of cover concrete was observed.

: Opening of Joints: Limit State ofConnecting Bolts

Lateraldisplacement

Figure 10 Limit states of precast hybrid concrete columns

Response acceleration (gal)1200
case1-80%
case2-80%800

加速度（gal）

400

0

-400

-800

-1200
-80

-60

-40

-20

0

20

40

60

80

変位（mm）
Lateral displacement(mm)

Figure 11 Seismic Response and Failure Mode after Design Ground Motion Level
Test in Post-repair Series for Precast Hybrid Column

129Limit States of Precast Reinforced Concrete Bridge Column
It was confirmed that the nonlinear behavior of the precast reinforced concrete
bridge column was similar or even better than the conventional reinforced concrete
bridge column because the longitudinal reinforcing bars inserted into the mortalgrouted ducts have better performance on anti-buckling. The limit states of this column
can be determined to be the same as those of the conventional reinforced concrete
columns, which is shown in Figure 7. Figure 12 shows the force-displacement
hysteresis and the failure mode obtained from the shake table tests during the designlevel earthquake ground motion. Only flexural cracks were observed, and the stablehysteresis loop was obtained.

Response acceleration (gal)
1500
1000
500
0
-200

-150

-100

-50

0

50

100

150

200

-500

150%

-1000
-1500
水平変位（mm)
Lateral displacement(mm)

Figure 12 Seismic Response and Failure Mode after Design Ground Motion Level
Test for Precast Reinforced Concrete Column
Conclusions
This paper introduces the state-of-practice of the conventional segmental
concrete bridge columns in Japan. Also, recent research works for the development of
the new precast segmental concrete columns were introduced in this paper. Design
Guidelines for the seismic effect of the precast segmental concrete columns isscheduled to be published in 2010. The seismic performance and the seismic limit
states of the precast segmental columns will be described in the guidelines based on
discussion of the test results. Furthermore, connection design details of those columns
will be specified.

Acknowledgments
Some illustrations and photos of the structural details of the precast segmental
concrete bridge columns in this paper were provided by PC-WELL Method
Association, Kajima Co., Sumitomo Mitsui Construction Co., Ltd., and P.S. Mitsubishi
Construction Co., Ltd. Author would like to appreciate their contribution for the paper.

130

Accelerated Constructions of the Viaductson the Second Keihan Expressway
Yoshihiko Taira1
Hirotsugu Mizuno2
Kei Muroda 3
Akio Kasuga4
Abstract
The Second Keihan Expressway located between Kyoto and Osaka has been
constructed as a bypass of the existing national route. Since the expressway islocated in the suburban residential area, cost saving, accelerated construction,
environmental protection for the surrounding residential area as well as improving
the safety were required for the construction of the expressway. To meet theserequirements, unique rationalized construction methods were applied for thedifferent site conditions in two viaduct projects, the viaducts in Nasu-dukuri Area
and in Aoyama Area.
This paper describes these new erection methods.
1. Introduction
The Second Keihan Expressway linking Kyoto and Osaka has been
constructed as a bypass of the existing national route (Fig.-1). Since theexpressway passes the suburban residential area, accelerated construction andreduction of the environmental impact during construction as well as improving
the safety etc. are required for the construction of the expressway. To meet theserequirements, two kinds of rationalized construction methods were newly
developed and applied in two viaducts in Nasu-dukuri Area and in Aoyama
Area, which are suitable for their own site conditions.
In Nasu-dukuri Area, since some degrees of construction area could be usedin the construction site, the areas were used as casting yard and a unique Ushaped precast girders were adopted. It was not as the conventional erection
method that uses muli-number of small precast segments divided longitudinally.
It was possible to have some casting yard along the viaduct and to transprt thesegment from the casting yard to the below area of the viaduct. The erection
method was called “U girder lifting erection” in which site-fabricated U-shaped
precast girders were used.
On the other hand in the viaduct in Aoyama Area, it was impossible to haveenough construction yard around the viaduct, and also it was impossible to use
the area under the superstructures due to the topographical conditions. The
segments were fabricated at the concrete factory and transported to the site. The
00000000000000000000000
1
Senior Structural Engineer, Sumitomo Mitsui Construction,
2
Manager, Naniwa National Highway Office, MLIT,
3
Senior Structural Engineer, SMC, 4Chief Engineer, SMC

131

girder was devided into several segments longitudinally and the alreadycompleted deck surface was used as the assembling yard of the segments. The
transported segments were lifted up, put on the deck and jointed as a girder.
Then the girder was transported on the deck to the newly erecting span with theerection girder. This erection method is called “Span-by-span erection with rearassembly system”.
In both viaducts, accelerated construction could be achieved by developing
the conventional erection methods, which are suitable for their own construction
conditions. It could shorten the erection cycle and could also save theconstruction cost.

Fig.-1 The Second Keihan Expressway,

2. Project summary
2.1 General features
General views of the viaducts are shown in Fig.-2 and Fig.-3, and the crosssections of the girders are shown in Fig.-4 and a Fig.-5, respectively. The
project summary and the viaduct properties are also shown in Table-1. The
configurations of these two viaducts such as width, total length, span length andthe structural type are quite similar. The design-built biddings were applied, andthe construction methods were proposed by the contractor.

Fig.-2 Nasu-dukuri ViaductFig.-3 Aoyama Viaduct132

29,640695

13780

29,640

810

13660

（下り線）

695

695

（上り線）

i=2.5%

13,780690（下り線）

i=2 .5%i =2.5%i= 2.5%

13,780

（上り線）

2,800

2,800

Fig.-4 Cross section of Nasu-dukuri ViaductFig.-5 Cross section of Aoyama Viaduct

Table-1 Project outlines and properties
Project

Nasu-dukuri Viaduct

Aoyama Via duc t

Profile(20 ,2@2,2@7)spans
pre stre ssed concrete viad uct

20 spans
prestressed concret e vi adu ct
Sept.2007-De c.20 09

PeriodMar.200 7-Mar.2009

Le ngth

79 0m

81 2m

Sp ans37m, 40.75m, 4 2m

40m, 41 m

Effective Width

2 @13.780m

13.780m,13.660m

Alignme nt
Verti cal Alignment
Horizontal Alignment

R=195 0-A=650-R=∞
1.240%-0.300 %2.5 00%

R=∞-A=5 00
2.96 6%-2.366%
2.500%

2.2 Requirements for the projects
During construction periods in both projects, following requirements were
imposed.
1) Each construction period is about two years. However, considering thetime for the detail design and other preparation work, only 18 months wereremained as the direct construction periods. Therefore, strong time reduction
was required.
2) Both viaducts are located in the quiet residential area, especially in
Aoyama Area. Therefore, the environmental impact had to be strongly reduced.
3. Construction of the viaduct in Nasu-dukuri Area
3.1 Outline of the erection method
In Nasu-dukuri they could use some degrees of construction yard at the siteand precast girders were selected. After fabrication the precast girder, the girder
was transported to the erected span and lifted using erection girder. The Ushaped girder with no upper slab was first applied in Furukawa viaduct on the
New Meishin Expressway (2002)1). This erection method is called “U girder
lifting erection method” (Fig.-6).
The procedures of the U girder lifting erection are shown in Fig.-7. Maximum
weight of a U girder is 2,400kN, and the girder was lifted with the erection
girder.
Since the precast girders were lifted near the supports of the erection girderon the piers, the bending moment acting on the erection girder could be quite
small. Furthermore, applying the U-shaped girder without upper slab could also
reduce the weight of the erection girder. As results, the bending moment acting
on the erection girder could be reduced up to 1/6 compared with theconventional span-by-span erection, and the erection girder could be lightened
drastically (Fig.-7, 8 and Table-2).
The following effects to the surrounding area could be expected.

133

695

1) Noises and vibrations caused by the construction work can be reduced
since the girders are fabricated at the fixed fabrication yard in the site.
2) No need to use the large trailers to pass the existing residential area fortransporting the precast segments.
The standard construction cycle of the superstructures is shown in Table-3.
In Nasu-dukuri Area, no stockyard was built for the U girder. This could beachieved by that the casting cycle of four girders were set to be equal to theirerection cycle. In order to achieve the cycle, four casting beds were used, andtwo sets of lifting girders were used in order to achieve the two weeks erection
cycle of four girders per span.
cast-in-s itu slab

Precast Panels

U-g irder

Fig.-6 Girders and slab structuresin Nasu-dukuri ViaductFig.-7 Conventional span-by-span erection
Cast-in-situ slab
Precast Panels
Lifti ng erection girder
U-shaped precast girder

Pier segment

Ｕ 形 断面 ﾌﾟﾚ ｷｬ
ｽﾄ 桁
U-shaped
girderpier segment

Fig.-8 Overview of the U-shaped girder lifting erection

134

Table-2
Erection Method

Erection methods and the bending moments of the erection girder

U-shaped girder lifting

Box girder lifting

Conventional span by span

U-girderBox girder

Box girder

4100 kNm
18%

6800 kNm
30%

23000 kNm
100%

Cross Section
at Erection

Max Bending Mo ment
in Erect ion g irder

Table-3 Construction cycle of Nasu-dukuri ViaductU-girder1
U-girder2

2

T

E
T

U-girder3

3

E

4

5

6

J

C

S

J

C

7

8

9

10

11

J

C

S

E

J

C

S

12

S
T

U-girder4erection
girder

1

E
T

Er

T : Tran sp o rt ati o n
Er

C : C u rin g

E : Erecti o n

S : St ressi n g

J : j oi nt co n cret in g

Er : Erect io n Gird er Eq ui pm en t

3.2 Fabrication and erection of the U girder
Considering the limited construction area and the way to transport theconstruction materials as well as the construction cycle work, four sets ofcasting bed were arranged in line longitudinally along the viaduct (Photo-1).
The U girders were transported from the casting bed to the erecting span witha large trailer (Photo-2). Photo-3 shows the erection of the U girder. Since large
tensile stresses occur near the lifting points inside the webs during lifting, thevertical prestressing and the additional reinforcement were arranged for the
local stress. Prior to the fabrication, a mockup test was conducted to confirm thestress and the safety during erection.
After lifting the girder, the girder was moved horizontally and transversally tothe fixed location, following that the U girder No.2 was lifted by anothererection girder. The U girder No.2 was then moved also horizontally andtransversally (Photo-4).
150 mm closure joints were placed at both ends of the girder and the piersegments. After placing concrete, each girder was externally prestressed andthen the lifting devices were released.
Segments No. 3 and No. 4 were also constructed as the same ways and theconstruction of one span was completed.

135

Photo-1 Casting yard

Photo-2 Transportation of the U-girder

Photo-3 Erection of the U-girder

Photo-4 Transversal movement of the girder

3.3 Construction of slab
For the construction of the slab, precast and prestressed panels were place on
the webs of the U girder. These panels are used as structural components as well
as for the formwork, and the re-bars were assembled on them. For the thermal
stress and the stress caused by the deformation of the girder due to the influenceof the concreting in the adjacent slab, as well as shrinkage and creep effects.
Expansive admixture was used as well as additional reinforcement designed
through the thermal stress analysis (Photo-5, 6).

Photo-5 Precast panels

Photo-6 Cast-in-situ slab136

4. Construction of the viaduct in Aoyama Area
4.1 Outline of the erection method
Since it was impossible to have the casting yard near the construction site andit was difficult to use the area below the superstructures in Aoyama Area,
existing concrete factory was utilized as the fabrication yard for the precast
segments of the girder. After the factory-fabricated precast segments weretransported from the factory to the construction site, the segments are then lifted
and assembled on the already constructed deck surface as an assembling yard.
This erection method is called “the span-by-span erection with rear assembly
system”. The core segment without some length of overhang slab could save theweight of the girder (Fig.-9). Compared with the viaduct in Nasu-dukuri Area,transversal movement devices were used in order to reduce the number of
erection girder, and the cost of the construction girder was reduced substantially.
Construction procedures of the erection method is shown in Fig.-10 asfollowed, and the standard construction cycle of the superstructures is shown in
Table-4.
1) Segments of the girder No.1 are put and jointed together on the alreadyconstructed deck surface. The assembled girder is transported toward theerection girder along the deck surface and erected with the erection girder.
The weight of a girder is about 3,500kN.
2) The girder No.1 is horizontally and transversally moved by the devices andthen tensioned. The following girder No.2 is also assembled, transported
and erected from the rear span continuously.
3) After the girder No.3 and the girder No.4 are erected in sequence and all
the four girders are erected, precast panels are placed and the erection
girder is moved toward the next span.
4) Slab concrete is placed.
In Aoyama Area, transporting the segments, lift up to the deck surface andjoining were performed in one day, and the transportation, erection andtransversal movement on the next day, while the segments of the next girder weretransported to the site in 2 days later. Joining work of the next girder was
performed concurrently at the rear assembling yard while the installation andtensioning were performed in the erecting span. As results, the construction cyclein one span of four girders took two weeks and was achieved as the same as thatin Nasu-dukuri Area.
Compared with the conventional span-by-span erection method, this erection
method with one set of erection girder could construct 1/2-1/3 faster by the
number of days. In other words, 2 to 3 sets of erection girder would be needed
to have the same erection speed in the conventional erection method (Fig.-11).

137

Cast-in-s itu slab

Precast panels

Core segment

Fig.-9 Girders and slab structures in Aoyama Viaduct

Assembling No.1 girder
Erection of No.1 girder
girderAssembling No.2 girder
Erection of No.2 girder

Transportation of No.2 girder

Precast panels

Cast-in-situ slab

Fig.-10 Overview of the span-by-span erection
with rear segment assembly system
Table-4 Construction cycle of Aoyama Viaduct1
main
girder１
main
girder2
main
girder3
main
girder4
erection
girder Er

2
E

3

4
J
E

5
S

6

7

8

9

10

11

J

S

Er

J

S

Er

12

Er
S

J
E

Er

J
E

Er

E : Erecti o n

J : j oi nt co n cret in g

S : S t res si n g

Er : Erecti o n Gi rd er Equ ip m ent

4.2 Assembly and erection of the segments
13 numbers of segments were transported from the factory with trailers. The
segments were placed on the deck surface with a crane, jointed together as agirder (Photo-7). Then the girder was prestressed and put on the carriers.
Devices for transversal movement were installed at the upper part of bothends of the girder, and the girder was transported toward the erecting span
(Photo-8). Since the weight of the girder and the devices reaches about 3,500kN

138

and they were transported on one of four existing girders, the stress of the girder
due to the loading was verified and additional prestressing tendons werearranged.
After the girder was transported toward the erecting span, the girder was hungby the crane installed on the erection girder (Photo-9). The devices fortransversal movement were installed on the rail on the pier segments, and thedevices with the girder were moved transversally to the fixed location (Photo10).
As the same as the viaduct in Nasu-dukuri Area, 150 mm closure joints wereplaced at both ends of the girder and the pier segments, supported with thetransversal movement devices. After placing concrete, each girder was
externally prestressed and then the supporting devices were released.

Photo-7 Assembling the segments

Photo-8 Transportation of the girder

Photo-9 Erection of the girder

Photo-10 Transversal movement of the girder139

4.3 Construction of slab
Precast panels were placed between the top of the girders (Photo-11), and rebars assembly and concrete placing were conducted (Photo-12).
By using precast panels with labor saving, the slab could be constructed at thesame days of the erection of four girders in next span. These procedures werequite effective ways to achieve the required construction cycles.

Photo-12 Cast-in-situ slab

Photo-11 Precast panels

5. Afterword
It has been considered that the conventional span-by-span erection using
conventional multi precast segments is suitable for the construction of the largescaled continuous urban viaduct project for the cost saving, achieving the highquality and for the accelerated construction. However, in the urban viaductprojects, there might be some severe site conditions as mentioned. In such cases,
the construction methods adopted newly developed in the viaducts in Nasudukuri Area and Aoyama Area on the Second Keihan Expressway can be thegood solutions in different site conditions. In both projects, the rate ofconstruction speed of 2,400m2 per month were both achieved.
Reference
1) Ikeda S., Ikeda H., Mizuguchi K., Muroda K., and Taira Y.: Design andConstruction of Furukawa Viaduct, Proceedings of the 1st fib Congress, Osaka,
2002

140

New FHWA Seismic Hazard Mitigation Studies for Highway Bridges

W. Phillip Yen, PhD, PE 1

Abstract
Earthquakes are inevitable events in our living environment. Each large
magnitude earthquake located around urban area created devastated destruction of ourinfrastructures, including the transportation system, and claimed vast human fatalities.
Since 1992, FHWA has initiated three major research projects in the seismic hazard
mitigation, they are Seismic Vulnerability Study for Existing and New Highway
Constructions, Seismic Vulnerability of Highway System, and the SAFETEA-LU
Seismic Research Program.
This paper describes the Federal Highway Administration’s (FHWA) new
seismic research program to mitigate earthquake loss of highway infrastructures. This
program consisted with two major research studies started in 2007 (five-year plan),the first is FHWA/MCEER project, titled, Innovative Technologies and Their
Applications to Enhance the Seismic Performance of Highway Bridges, and thesecond one is FHWA/ UNR project, titled, Improving the Seismic Resilience of theFederal-Aid Highway System.

Introduction
Surface transportation is a vital component of our society. Our highways link
airports, train stations, harbors, manufacturing plants, farms office and residences.
This transportation network must continue functioning during and after a natural
hazard such as an n earthquake so that the lifelines of our society may be restored assoon as possible. Of all the components of the surface transportation system bridges
are the most vulnerable to earthquake damage.
About 65 percent of the approximate 600,000 bridges in the U.S. wereconstructed prior to 1971 with little or no consideration given to seismic resistance.
Recent earthquakes such as Loma Prieta, CA in 1989, Northridge, CA in 1994 andKobe, Japan in 1995 & Chi-Chi Earthquake, Taiwan in 1999, have demonstrated theneed to find new ways to build earthquake-resistant bridges and highways, and to
retrofit existing bridges.
Recognizing the shortcomings evident in both existing bridges and design1

Program Manager, Seismic Hazard Mitigation, Office of Infrastructure, R&D Federal Highway
Administration, 6300 Georgetown Pike, McLean, VA 20121 Email address:
Wen-huei.Yen@fhwa.dot.gov

141specifications, the Federal Highway Administration (FHWA) initiated several
comprehensive seismic research studies on bridges and highways since 1992. As aresult, a new seismic retrofitting manual, consisting of two parts (bridges and otherhighway structures), was completed and published. A new comprehensive seismic
design recommendation was also published under the seismic vulnerability studies.
Furthermore, FHWA also published the research products in seismic
retrofitting of truss bridges, seismic isolation manual and Risks from Earthquake
Damage to Roadway System (REDARS). To implement these technologies into ourFederal-Aid transportation system with limited resource, cost-effective and practicalmethods such as accelerated bridge construction, are also essential to successfully
improve the seismic safety.
Under the new Transportation Authorization SAFETEA-LU, FHWA is
working with Multi-disciplinary Center of Earthquake Engineering Research(MCEER) of New York State University at Buffalo and University of Nevada at Reno
(UNR) to initiate two major seismic research studies and started in 2007. The
following are the summary of these two new studies:

Seismic Research Study 1 (with MCEER): The Innovative Technologies andTheir Applications to Enhance the Seismic Performance of Highway Bridges
The objective of this study is to improve the seismic resistance of our highway
system, by developing new innovative technologies and their applications, bydeveloping cost-effective methods for implementing design and retrofitting
technologies, and by refining and expanding applicability.
This project is to increase the mobility and safety of our surface transportation
system as the FHWA envisions reducing the construction/ maintenance time of new
and existing highway structures. Applying accelerated bridge construction
technology to high seismicity area requires more advanced connection detail toaccommodate the large ground motions. Innovative technologies and theirapplications are continuously sought to refine and expand their applicability toenhance the seismic performance of our surface transportation system.
The major tasks of this study are:Developing Detailed Technology to Apply Accelerated Bridge Construction (ABC) in
Seismic Regions
This task is to develop implementable seismic design guidelines applicable tothe bridges located in the seismic regions. The focus of this task is on prefabricatedreinforced concrete, segmentally constructed highway bridges of short to medium
span length. A technical monograph for the bridge system is expected from this study.
In addition, a separate subtask is set up to develop recommended design guidelines
with design examples for practical applications.

142

Innovative Seismic Protection Technologies
In this task, innovative technologies will be explored that can enhance the seismic
performances of precast R.C. Bridges with emphasis given to ABC. This task will
include subtask study on:






Design Guidelines and Demonstration of Roller Isolation Bearings,
Lifetime Performance of Bridges with Seismic Protective System,
Bridge Information Modeling for Seismic Aspects of Accelerated Bridge
Engineering
Development of Structural Fuse Concept for Bridges
And Down-Scaled Bridge Pier Testing

Seismic Study 2 (with UNR): Improving the Seismic Resilience of the FederalAid Highway System

BackgroundToday life-safety is no longer the sole requirement for the successful design ofa highway system for a major earthquake. Resilience is now expected by the travelingpublic as an integral component of any design strategy, so as to ensure rapid recovery
and minimal impact on the socio-economic fabric of modern society. This realization
has led to the concept of performance-based seismic design which is a relatively new
development in the design and construction of civil infrastructure. Neverthelesssubstantial progress has been made in this area, particularly with respect to theperformance of individual components of the built environment, such as buildings andbridges. But the real potential for performance-based design comes when theseconcepts are applied to systems and subsystems of the infrastructure, such astransportation networks, subject to both service load conditions and extreme events.
Since 1993, the FHWA has been researching methodologies for seismic risk
analysis (SRA) under its seismic research program. The result of this effort has been
the development of an earthquake loss estimation software tool called REDARS,
which is now being used in pilot projects by several State Departments ofTransportation. REDARS has been specifically developed for assessing theperformance of highway systems taking into account the inter-connectedness of the
network and vulnerability of bridges to seismic loads.
Performance measures calculated by REDARS include congestion and delay
times. These measures allow system-level performance criteria to be specified forearthquakes of various sizes, such as maximum permissible traffic delay times andminimum restoration times. Accordingly the resilience of a highway system may be
defined and measured in quantitative terms, such as the time it takes to restore thesystem’s pre-earthquake capacity, as illustrated in figure 1. In doing so, financial and143societal incentives can be developed that will improve resilience and at the same time
reduce risk to life and property.
Q
Resilience
Q0
Vulnerability
Recovery
Q1

t1

t2

t

Figure 1. System resilience, after Bruneau(Buckle and Lee, 2006).

Whereas REDARS is the result of a decade-long period of development, andrecently shown to adequately replicate the performance of the highway system in the
San Fernando Valley following the 1995 Northridge earthquake, there is still much to
be done to enable the methodology to be used with confidence and be widely
applicable. REDARS has been developed with the expectation that new and moresophisticated modules will be developed overtime, in order to improve its accuracy
and expand its range of application. This is considered a critical step in the drive
towards quantifying the resilience of the highway system.
For example, lessons learned from recent large earthquakes, such as the 1999
Chi-Chi earthquake in Taiwan, have indicated the importance near-fault effects on
bridge response. At the FHWA/NCEER Workshop on the National Representation ofSeismic Ground Motion for New and Existing Highway Facilities held in San
Francisco in May 1997, a consensus was reached that a response spectrum alone is
not an adequate representation of near-fault ground motion characteristics, because it
does not adequately represent the demand for the high rate of energy absorption
imposed by near-fault pulses. This is especially true for high ground motion levels
that drive structures into the non-linear range, invalidating linear elastic assumptions
on which the elastic response spectrum is based. Near-fault ground motions aredifferent from ordinary ground motions in that they often contain strong coherent
dynamic long period pulses and permanent ground displacements. The dynamic
motions are dominated by a large long period pulse of motion that occurs on thehorizontal component perpendicular to the strike of the fault, caused by rupture
directivity effects. Near fault recordings from recent earthquakes, such as Chi-Chi
earthquake, indicate that this pulse is a narrow band pulse whose period increases
with magnitude, as expected from analysis.
Many cities on the West Coast of the U.S. are located in near-fault
environments and this should be taken into account when studying the resilience of
highway systems on the West Coast in particular. Thus bridge fragility functions are
required that include near-fault effects for inclusion in loss-estimation models such as
REDARS. In addition practical recommendations are required for the design ofhighway bridges subject to near-fault effects.

144

The objective of this project is to study the resilience of highway systems witha view to improving the performance of these systems subject to major earthquakes. A
comprehensive assessment tool to measure highway resilience shall be developed by
improving current loss estimation technologies, such as REDARS; factors affecting
system resilience will be identified such as damage-tolerant bridge structures andnetwork redundancy; design aids for curved bridges and those structures in near-fault
regions will be developed; new technologies will be developed for improving theseismic performance of bridges; methodologies and technologies developed herein
will be implemented in REDARS to the extent practical; and outreach to improveseismic safety will be conducted.
The following are the major tasks of this study:
REDARS Customization for Resilience Studies
This task shall implement those upgrades for resilience studies that will focus
on improving computation and display of important parameters for characterizing
resilience.
Characterizations of Seismic Hazards for Near-Fault Bridges
This task will work on Ground Motions including Effects of Rupture
Directivity and Surface Fault Rupture Hazards to improve its characterization of
surface fault-rupture displacements with a focus on faults as a single straight line.
Seismic Response of Horizontally-Curved Highway Bridges
This task is to perform a comprehensive study of the seismic response ofhorizontally-curved highway bridges. A set of seismic design guidelines for this typeof bridge shall be developed.
Near-Fault Bridge Study
This task is to develop recommendations for procedures that can be readily
used to design bridges in the vicinity of faults and improve public safety throughout
the United States.
Fragility Function for Curved, Near-Fault and other Bridges.
This task will focus on the development of fragility functions for both curved
bridge and near-fault bridges, and other bridges.
Concluding Remarks
Risk mitigation methods to reduce earthquake losses need a great effort fordevelopment and implementation. The most difficulty with mitigating earthquake
hazards is that earthquakes come without any notice. There is no way to accurately
predict when an earthquake will occur, nor what its magnitude will be. Earthquakes

145are devastating, often resulting in a great number of deaths, injuries and extensive
infrastructure damage. Losses will occur in just one or two minutes. Systematic
approaches to evaluating earthquake risks, including direct and indirect losses such as
economic impact, have become an important issue in our engineering community.
The above two studies focus on (1) The Innovative Technologies and Their
Applications to Enhance the Seismic Performance, and (2) Improving the Seismic
Resilience of the Federal-Aid Highway System. These two studies will produce many
practical recommendations for bridge design in accelerated bridge construction arena,
and advance the current design guidelines of curved and near-fault bridges. The study
will enhance the previously developed REDARS program with more meaningful andaccurate fragility functions to estimate the earthquake losses.

Reference:
FHWA Congressional Seismic Research Studies under the SAFETEA-LU Program.

146

OVERVIEW OF THE DEVELOPMENT PROCESS AND EFFECTS OF
THE UFO METHOD(ACCELERATED CONSTRUCTION METHOD FOR OVERPASSES)
Yuji Mishima1, Yukio Katsuta2, and Tomoaki Tsuji3
Abstract
Traffic congestion at intersections is a serious problem in urban Japan because it
causes economic loss and environmental degradation. Shorter construction periods
and minimal traffic restrictions are needed to prevent further congestion during theconstruction of overpasses. The Uni-Fly-Over (UFO) method was developed toachieve dramatically shorter construction periods through structural rationalization
for foundations. This paper describes projects for which the UFO method has beenused to reduce construction time, explains the method's advantages such as its
mitigation of environmental degradation, and outlines a review on the technical issues
that had to be overcome to develop steel spread foundations.
Introduction
According to a recent survey conducted by the Ministry of Land, Infrastructure,
Transport and Tourism (MLIT), there are over 2,000 intersections throughout Japanthat are in need of a solution to the problem of traffic congestion. This indicates thatthe mitigation of traffic congestion at intersections and railroad crossings is a key
issue in realizing an improvement in urban road functions.
Although overpass of intersections is an extremely effective technique foralleviating traffic congestion, the construction of a grade-separated junction in areas
suffering from congestion creates further congestion due to the traffic restrictions thatmust be enforced during the work or aggravates the roadside environment as a result of
the noise and vibrations generated in the course of the work.
Given this, there is a growing social need for an accelerated construction
technology that will reduce the economic loss that results from traffic restrictions andavoid environmental deterioration by shortening the construction period for
overpasses.
The UFO method was developed in 1989 by Hitachi Zosen Corporation to provide
an engineering solution to society’s demands for a technology that would accelerate theconstruction of an overpass with a spread foundation. This method has been applied in
the construction of three bridges to date. The greatest advantage of this method is theconsiderable reduction in local construction time it offers by enabling the use oflightweight, prefabricated steel members in the construction of the foundations, which
has been a very time-consuming task until now.
This paper describes the development concept and characteristics of the UFO
1Manager, Infrastructure Technology Development Dept., Hitachi Zosen
Corporation, Japan
2
Manager, Bridge Construction Dept., Hitachi Zosen Corporation, Japan
3
Assistant Manager, Bridge Design Dept., Hitachi Zosen Corporation, Japan147

method and explains its advantages by looking at actual cases in which the method
has been applied. The paper also outlines a review made of the technical problems
involved in the development of the steel spread foundation, one of the characteristicfeatures of the UFO method, based on some case studies.
Development Concept
The principal purpose of developing the UFO method was to reduce the local
construction time for overpasses—including the construction of the superstructure,
substructure, and foundations—and to minimize the period of time for which traffic
restrictions are imposed. Its development involved the study of conventional
problems and their potential solutions. Once that had been done, a method thatsatisfied the following conditions was realized:
1)
2)
3)

4)
5)

Earthwork and concreting work, which are both time-consuming activities
performed at the work site, should be minimized.
Members should be prefabricated and a structure that reduces theconstruction time at the work site should be used.
Structural rationalization and downsizing should be carried out with respect
to the foundation structure, whose construction takes place mainly at thework site and is very time-consuming.
The retaining wall at the approach should be constructed at the same time as
the overpass.
The members should be erected using readily available machines, andmembers that can be erected without the use of special equipment should be
used to ensure the universality of the method.

Originally intended for use in the construction of spread foundations that are laidon the bearing layer about 5 m underground, the UFO method was developed tosatisfy the above-mentioned five conditions. The aim was to achieve a substantial
reduction in the local construction time of over 50% compared with conventional
methods. An accelerated construction method for pile foundations that was intended
for deeper bearing layer was developed jointly by PWRI and Hitachi Zosen. Knownas the Uni-Anchor System, it provides a rational combination of steel pier and pilefoundations, as shown in Fig. 1. (PWRI. 2005; Mishima et al., 2005)
Characteristics of UFO Method
Fig. 2 provides a structural overview of the UFO method that was developed
based on the above-mentioned concept. The structural features are as follows:1)

Continuous steel rigid-frame box-girder bridge with steel deck andintegrated superstructure and substructure
To facilitate quicker construction work, the superstructure and substructure areconstructed using prefabricated steel piers and box girders with steel deck. An
integrated rigid-frame structure is employed for the superstructure and substructure to148

provide enhanced earthquake resistance. As this structure also eliminates the need forbearings and expansion joints, it also provides a smoother ride for vehicles and saving
in maintenance costs.
2)

Steel spread foundations
As the superstructure and substructure of a conventional overpass are made of aconcrete structure with a substantial dead load, larger foundations are inevitably
required. This not only necessitates an extension of the work period, but also anenlargement of the work yard, which ultimately means that strict traffic restrictions
have to be imposed for the roads at and around the construction site.
The UFO method offers a reduction in weight because the use of steel members
for the superstructure and substructure results in the foundations being subjected to asmaller load, enabling prefabricated steel members to be used for the foundation
structure. More specifically, a foundation structure built using the UFO method
consists of steel members—namely, supporting beams and connecting beam—laid out
in a grid pattern with a thin reinforced concrete slab laid underneath them, as shownin Fig. 2. The load from the superstructure and substructure is efficiently dispersed
into the ground from the steel members arranged in a grid pattern via the footing. Fig.
3 compares the reaction dispersion mechanism for the UFO method with that of aconventional method. As shown in this figure, the spread foundation is divided into
two types: continuous footing along the bridge axis (hereinafter, “continuous
footing”), which is a series of steel footings installed between piers along the bridge
axis; and isolated footing along the bridge axis (hereinafter, “isolated footing”), which
is a single steel footing installed for each bridge pier along the bridge axis.
Continuous footing is generally employed when the bearing capacity of ground isrelatively low and a large foundation would be required if isolated footing were used.
It is also used if it is necessary to disperse the subgrade reaction into the ground as aresult of the interaction between the foundations and any underground buried objects.
On the other hand, isolated footing is generally employed if any underground buried
objects present has no impact on the foundations and the bearing capacity of groundis relatively high.
As mentioned earlier, the UFO method makes use of prefabricated members foralmost all the members of both the superstructure and substructure and compact
members for the foundation structure. It therefore realizes a significant reduction in
the work period and the size of the work yard compared with conventional methods.
3)

On-site construction method
The basic method for erecting steel members involves the use of truck cranes,
which are economic, versatile, and highly mobile. Use of this erection method
enables the following to be achieved: a reduction in the size of the work yard, thecyclic erection of members using the dividing strip as the work yard, and thesimultaneous construction of approach section (retaining wall).
The Harada Viaduct is shown in Fig. 4 as an example of the standard procedure
employed with the UFO method. As the figure shows, the procedure involves groundexcavation for installation of a pier foundation and construction of the concrete slab.
This is followed by the cyclic erection of the foundation members, bridge piers, and149

the superstructure, in this order. After the erection of the side spans has been
completed, the central spans are erected in the direction of the intersection, andfinally upper spans are set in place directly above the intersection. Since the degree ofprecision achieved in the fabrication of the steel members has a great impact on the
degree of precision achieved in their erection, the finished quality of the members
needs to be strictly controlled by temporary assembly in a fabrication shop, as shownin Fig. 5.
UFO Method Case Studies
1)

Kitahanada Overpass
Conditions at the construction site for the Kitahanada Intersection (located in
Sakai city, Osaka prefecture) were extremely challenging because the area in which it
is located is home to a major intersection and the Kitahanada subway station. To
achieve an uninterrupted traffic flow at this junction, Osaka Prefecture began a
project to construct what would become known as the Kitahanada Overpass in 1994,
and construction work was completed in 1996. The UFO method was employed forthis project so as to realize a considerable reduction in construction time andminimize the impact of the overpass on the subway station structure.
The overpass is a 15-span continuous steel rigid-frame box-girder bridge withsteel deck, as shown in Fig. 6. When the structural composition of the overpass was
first planned, a design had to devised that would ensure that the bridge construction
would have no impact on the subway station structure. This was necessary because
when the subway station was constructed, no consideration had been given to thepossible addition of any extra load. Specifically, the steel spread foundation wasburied underground and lightweight EPS (expanded polystyrene) was used as backfill
to make the weight of the excavated soil equal to the weight of the bridge structure.
To ensure an even dispersion of the subgrade reaction, a continuous footing was also
employed for the steel spread foundation, based on which the bridge pier intervals
were determined. In addition, the soil under the bridge piers was improved to achieve
uniform bearing capacity of ground.
Since this was the first time that a foundation structure had ever been built using
the UFO method, a 1:10 scale model, as shown in Fig. 7, was produced to conductloading tests. These tests confirmed its safety with respect to the subway stationbuilding.
Thanks to the UFO method, the superstructure and substructure were completed
in just 12 months. It was estimated that had a conventional method been taken, it
would have taken 27 months to build a rigid-frame bridge comprising 2-spancontinuous non-composite steel I-girder and 3-span continuous steel box girder withsteel deck as the superstructure and a pile foundation as the substructure. This meansthat a reduction ratio of over 50% was achieved. (Hayakawa et al., 1997)
2)

Harada Viaduct
The Shikoku Regional Development Bureau of the MLIT decided to employ anaccelerated construction method in the construction of a viaduct over the Harada

150

intersection (Marugame City, Kagawa Prefecture) to mitigate traffic congestion andprevent traffic accidents. In selecting the construction method during the developmentof the basic plan, the Shikoku Regional Development Bureau chose to review 12
accelerated construction methods that might be suitable, comparing them on the basisof construction period, cost, and constructability. Following this review, the UFO
method was adopted as the standard method because it had a proven track record andreceived high evaluations overall. Tenders were invited in 2003 for the constructionof a viaduct over the Harada intersection, and the tenderers were screened using thecomprehensive evaluation method for which the main evaluation items were theconstruction fee and how few days the project could be completed in. The project was
successfully completed in 2004. Fig. 8 shows the erection work and the completed
viaduct.
The viaduct is a 7-span continuous steel rigid-frame box-girder bridge with steel
deck, as shown in Fig. 9. Because a bearing layer with an N value of 30 or more wasfound at a depth of 3 to 8 m from the ground surface of the site, it was possible to use
fewer bridge piers in the plan for the viaduct’s structural composition than wererequired for the Kitahanada Overpass, which had extremely challenging groundconditions. Soil improvement was also conducted to reduce the amount of earth thatwould need to be excavated, which eventually helped to minimize the number of onsite construction days required. Continuous footing was used since a uniform
distribution of reaction in the improved soil was needed for the steel spreadfoundation.
With construction of the superstructure and substructure completed in just 250
days, the UFO method achieved a 60% reduction in the on-site construction period, as
shown in Table 1. In contrast, it was estimated that had a conventional construction
method been used, it would have taken 630 days to construct a hollow PC slab bridge
and a simple composite steel I-girder bridge as the superstructure and an RC bridge
pier and RC spread foundation as the substructure and foundation. This reduction in
the construction period meant that the method was also able to reduce the traffic
restriction period by 40%.
To assess the economic efficiency of the method, the time lost as a result ofcongestion during the traffic restriction period and the time gained as a result of theearly opening of the roads were calculated in monetary terms, the results of whichwere then evaluated together with the construction cost (see Fig. 10). This revealed
that using the UFO method is about 20% more expensive than the conventional
method, but about 2% more advantageous in terms of a comprehensive evaluation oflosses arising from congestion and gains from the early opening. Further advantages
are known to be gained from the use of the UFO method, including the prevention of
the environmental degradation, which cannot be valued in monetary terms and areduction in losses from disruption to goods distribution. (Yamada et al., 2005)
3) Tonyamachi Overpass
In 2007, the Kanto Regional Development Bureau of the MLIT invited tenders
for a project to construct a viaduct over the Tonya intersection (located in Utsunomiya
City, Tochigi Prefecture) to mitigate traffic congestion at the intersection point.

151

Executed as a package contract for the design and construction of the overpass, theproject involved large-scale construction work over a length of road totaling 1,190 m
in length that included the construction of an elevated bridge and approach as well as
road improvements to the section under the intersection. A joint venture in which we,
Hitachi Zosen Corporation, participated won the contract because our proposal
offered an on-site construction period that would, thanks to the use of the UFO
method for the construction of the overpass, be 50% shorter than the employer’s
requirement and because our bid price was the lowest of the three joint ventures who
participated in the tender. The project is currently underway, with completion
scheduled for March 2010.
The overpass is a 3-span continuous steel rigid-frame box-girder bridge with steel
deck as shown in Fig. 11. The dimensions of the bridge and the retaining wall were
determined in such a way as to achieve a good balance between the construction costand a reduction in the construction period. In addition, isolated footing, rather thancontinuous footing, was used for the steel spread foundation because, unlike for the
Kitahanada Overpass or the Harada Viaduct, underground buried objects had no impact
on the work, the bridge length was short, and the supporting ground had sufficientstrength. This rationalization allowed us to plan for a reduction in the standard on-site
construction period set by the employer of over 50%.
Technical Review of Steel Spread Foundations
1)

OutlineThe greatest advantage of the UFO method is its ability to significantly reduce
the on-site construction period and the size of the work yard by employing a steel
spread foundation composed of prefabricated members that are both lightweight andcompact. The use of such a steel spread foundation was unprecedented, and thegreatest challenge posed by the application of this method was to ensure its safety.
There are two types of steel spread foundation used in the UFO method:
continuous footing and isolated footing. Which of these methods is employeddepends on the bearing capacity of ground and the conditions of any undergroundburied objects, as earlier mentioned.
Reviews conducted of these two types of footing are outlined in the following
sections using the Harada Viaduct and Tonyamachi Overpass Bridge as case studies,with a particular focus placed on safety.
2)

Continuous footing
If a spread foundation is being considered for use as a rigid foundation, as isconventionally done for a reinforced concrete footing, its safety should be verified
against the Specifications for Highway Bridges (Japan Road Association, 2002), andthe footing should then be designed to be thick enough to be verified as a rigid entity inaccordance with the said specifications. If a footing is confirmed to be thick enough as
a rigid body, this means that the equations for calculating the bearing capacity ofground or subgrade reaction as specified by the Specifications for Highway Bridges are
determined based on the assumption that the footing in question is a rigid body. The
type of spread foundation used for the Kitahanada Overpass and the Harada Viaduct152

was a continuous footing both along the bridge axis and in the direction perpendicular
to the bridge axis. For the direction perpendicular to the bridge axis, the rigidity of thefoundation as a rigid body can be maintained because of a shorter footing span. For the
direction along the bridge axis, however, since the footing span is 14 to 21 m, it is notrational either in terms of structure or construction for the footing to have sufficient
thickness as a rigid body. Therefore, it was necessary to consider the continuous footing
as an elastic foundation and to check the stability in terms of bearing capacity, toppling,and sliding by taking into account the uneven distribution of the reaction from thesuperstructure and substructure. Typical safety check methods include loading tests
using a scale model, as in the case of the Kitahanada Overpass, and finite element
analysis. The finite element analysis used for the Harada Viaduct is explained below toserve as an example.
For the Harada Viaduct project, both the local ground and the steel foundation
members determined relative to external forces during the design stage were modeled
for review using three-dimensional finite element analysis, which can accurately
consider the impact of the three-dimensional ground supporting the spreadfoundation, as shown in Fig. 12. For a conventional type of spread foundation, only atest load equivalent to the load that would occur in the event of a Level 1 earthquake
is used due to the behavioral characteristics of the foundations in the event of a major
earthquake. But since no behavioral characteristics for a steel spread foundation havebeen identified yet, the load that would occur in the event of a Level 2 earthquake
was used for the safety check to err on the side of caution.
The distribution of the level of vertical stress on the ground is shown in Fig. 13
as an example of the analytical results. This figure indicates that no stress in excess of
the allowable value was exerted on the ground and that stress was generally dispersed
across all of the ground at a depth of 6 to 7 m. The results for toppling and sliding
also revealed no problems.
In conclusion, the safety of steel spread foundations was verified using threedimensional finite element analysis, which takes into account the three-dimensionality of theground.
3)Isolated footing
If the ground strength is relatively high and underground buried objects do notcause any major problems, isolated footing that is compact and easy to constructserves best, as was the case with the Tonyamachi Overpass (Fig. 11). However, for an
isolated footing, steel spread foundation members are cantilevered along the bridgeaxis, which eventually causes a concentration of the reaction just under the bridge
pier if the footing rigidity is low. Therefore, it is necessary to provide steel memberswith a level of rigidity that is as high as that of a conventional spread foundation sothat the reaction force of the superstructure and substructure acts uniformly on the
ground.
As has already been mentioned, the required rigidity for a spread foundation is
specified in the Specifications for Highway Bridges. However, the provisions in these
specifications are meant for conventional spread foundations, which are made of153

reinforced concrete of high rigidity. As such these specifications cannot be applied
directly for lightweight, relatively flexible foundations such as foundations made ofsteel members and thin concrete slabs, as used in the UFO method. For a steel spreadfoundation, therefore, it was necessary to verify whether the steel foundation
members, whose cross section had been determined in the design stage according to
the external forces to which they would be subjected, possessed the required rigidity.
As a result, three-dimensional finite element analysis was used to verify this, as
was the case with the Harada Viaduct. Given that the subject for analysis was an
isolated footing, a partial model was used, as shown in Fig. 14. To be on the safe side,
the loads to be checked included those that would occur in the event of Level 2
earthquake. The section forces of the pier base, which were calculated from aseparately conducted dynamic analysis, were applied to the end of the model as aboundary section force.
The rigidity of the steel spread foundation was evaluated based on an existing
research paper (Iijima, 1981) that served as the basis for the rigidity judgment adopted
in the Specifications for Highway Bridges. More specifically, attention was paid to thelevel of subgrade reaction intensity, and the difference between the subgrade reaction
intensity (Pmean) for a spread foundation as a rigid body and the maximum subgrade
reaction intensity (Pmax) for a steel spread foundation was expressed as a ratio to Pmean(see Fig. 15). This ratio is taken as the dispersivity of reaction. If the allowable level ofthis dispersivity is within ±20%, the foundation should be taken as a rigid body.
As an example of this result, the distribution of the subgrade reaction in the eventof a Level 2 earthquake in the direction perpendicular to the bridge axis is shown in
Fig. 16. As the figure shows, the dispersivity of the subgrade reaction is within theallowable limits. This confirms that steel spread foundations have sufficient rigidity.
Highly precise results may be obtained from a rigidity evaluation of a steel
spread foundation if finite element analysis is used, but considering the time andmoney it would take to carry out such an analysis every time one was necessary, asimplified evaluation equation would be of great benefit for the actual design work.
Conclusion
This paper presents an outline of how the UFO method, an accelerated
construction method used to mitigate congestion at intersection points, was developedand introduces some projects for which the method was applied. The projects forwhich the UFO method was applied turned out to be successful in terms of theintended goal of achieving a significant reduction in the construction period, despite
differences in their on-site conditions.
The following points summarize the conclusions of this paper.
1)

The UFO method can be successfully used to reduce the on-site construction
period by over 50% compared with conventional methods, although theactual reduction varies according to the local conditions.
2) A reduction in the on-site construction period will produce various otheradvantages, including a reduction in the traffic restriction period and154mitigation of the negative impact on the environment.
3) The UFO method uses two types of spread foundation: continuous footing,which is a series of steel footings along the bridge axis, and isolated footing,which is a single steel footing along the bridge axis. A significant reductionin the on-site construction period can be achieved by selecting the type offoundation suited to the bearing capacity of ground and the conditions of any
underground buried objects.
4) Loading tests with a scale model or 3D FE analysis with a local groundmodel have verified that steel spread foundations are safe in the event of a
Level 2 earthquake.
References
Hayakawa, Kitaya, Yamashita, Beppu, Shiomi, and Wakita: “Design of theKitahanada Overpass—Built using a Spread Foundation Immediately Above the
Subway Station Building,” Bridge and Foundation Engineering, Vol. 31, No. 1,
Jan. 1997. In Japanese.
Iijima: “Footing Thickness,” Bridge and Foundation Engineering, Vol. 15, No. 5,
1981. In Japanese.
Japan Road Association: “Specifications for Highway Bridges and their
Explanation,” March 2002
Mishima, Tahara, Wakabayashi, Sasaya, Fukui, and Takeguchi: “Experimental Studyof Connection between Steel Bridge Pier and Foundation,” Bridge andFoundation Engineering, Vol. 41, No. 11, Nov. 2007. In Japanese.
Public Works Research Institute, Hitachi Zosen Corporation, and Fujita Corporation:“Joint Research Report on the Development of a Technique to Reduce the Onroad Construction Period for an Overpass Project,” Joint Research Report No.
334, March 2005. In Japanese.
Yamada, Toyosaki, Takeuchi, Mishima, Tahara and Ishiyama: “Design andConstruction of the Harada Viaduct—Overpass of an Intersection using an
Accelerated Construction Method,” Bridge and Foundation Engineering, Vol. 39,
No. 9, Sept. 2005. In Japanese.

[Uni-Anchor System (UAS)]
Steel Pier
Steel Shell

External
Steel Plate

Perforated Plate(PBL)

Cast-in-place Pile

Fig. 1 Structural Overview of the Uni-Anchor System

155

Steel Deck

Rigid-frame Box-girder Bridgewith Steel Deck
Main Girder
Supporting Beam

Steel Pier
Connecting Beam

Steel Spread Foundation
RC SlabFig. 2 Structural Overview of the UFO Method
Conventional Method
RC Pier

RC Members
Used→Dead Load:
Large

Reaction

RC Girder
RC Footing

Box-girder Bridge with Steel Deck

Steel Pier

UFO Method

Steel Spread Foundation

Box-girder Bridge with Steel Deck

Reaction

Steel Members
Used→Dead Load:
Small

Reaction Force
Concentrated Immediately
under Pier

Reaction

Steel Pier

Supporting Beams andConnectors Disperse
Reaction

Steel Spread
Foundation
Continuous Footing

Isolated Footing

Fig. 3 Comparison of Reaction Dispersion Mechanisms
STEP-1
Excavation and RC Slab WorkSTEP-1
掘削工、RCフーチング工Approach
Terre
Armee
テールアルメApproach
テールアルメ
RC
SlabRC
footing
RCフーチング

RC Slab
RCフーチング

A1

STEP-2

P1

P2

P3

P4

Steel
鋼矢板 Sheet Pile

RC
Slab
RC
footing
RCフーチング

P5

P6

START
START

Steel
Sheet Pile and
鋼矢板，掘削Excavation

A2

Erection of Superstructure and Substructure (Side Spans)

STEP-2 上下部工の架設（側径間部）
Erection
Direction
架設方向

RC Slab Work
RCフーチング工

Erection
Direction
架設方向
Erectionof Side Spans
側径間部の架設(Spread(直接基礎→橋脚→主桁,etc.)
Foundation→Pier→Main Girder, etc.)
ベント設備
Vent
Equipment

ベント設備
Vent Equipment
A1

P1

P2

P3

P4

P5

P6

160 t Hydraulic Crane
160ｔ吊油圧クレーン

160
t Hydraulic Crane
160ｔ吊油圧クレーン

A2
160ｔ吊油圧クレーン
160 t Hydraulic Crane

STEP-3上部工の架設（交差点部）
Erection of Superstructure (Intersection)
STEP-3

Erectionof Intersection
交差点部の架設(Main
Girder, etc.)（主桁，etc.）

Filled-in
and Encased Concrete Work
中埋め・根巻きコンクリート

閉合ClosureBackfill埋戻し，橋面工
and Bridge Deck Work

A1

P1

P2

P3

P4

160ｔ吊油圧クレーン
160 t Hydraulic
Crane

P5

P6

A2

END
END

Vent
Equipment
ベント設備

Fig. 4 Standard Construction Procedure (Harada Viaduct)

156

Fig. 5 Temporary Assembly for a Steel Spread Foundation (Harada Viaduct)
Length of Overpass
Section 415000 (on
CL)
立体区間長
(CL上)
Length of Approach 72000
Length of Approach 92000
Length of Bridge Section(Bridge Length) 251000 (CL上)取合部施工区間長(on CL)
取合部施工区間長
橋梁区間長(橋長)2500
2500
GirderLength(on
CL)
桁長250800(CL上)
30000
20000 49500
100
100
59500
2x14000=28000
L型擁壁
L-shaped
L-shaped Gravity retaining
L型擁壁(on CL) 2x14000=28000
Gravity
支間長
50000(CL上)
13950
13950
4x14500=58000
4x14500=58000
重力式擁壁Retaining
Span Length
retaining
重力式擁壁
wall
450
Retaining450
Box-girder
Bridge with Steel Deck Steel Spread Foundation
Wall
鋼床版箱桁
wall
Steel鋼製直接基礎
Spread Foundation
鋼製直接基礎
Wall

Kitahanada
Subway Station Caisson
地下鉄北花田駅舎函体

101100
A1

P1

P2

P3

P4

P5

P6

P7

鋼製橋脚
Steel Pier

101100P8P9P10 P11 P12 P13 P14

RC SlabSteel Spread
Foundation(Continuous
Footing)

Soil
Improvement

Subway Station
Structure

A2

Fig. 6 Structural Overview of the Kitahanada Overpass

Fig. 7 Loading Test with a Scale Model

157

Steel Spread Foundation
Full View of the Completed Viaduct(Continuous Footing)Fig. 8 Erection Work and Full View of the Harada Viaduct
側面

図Length of Viaduct
Section(on CL)
立体交差区間長
860 000 (CL上)

A1

5615

6450

5450

5450

7700

P1

P2

P3

P4

Soil
地盤改良

4843

Steel Pier
鋼製橋脚

4500
5376

32950

4970
1800
5396

Wall
重力式擁壁
600
133 900 63 900
ﾃｰﾙｱﾙﾒTerre
Armee

Lengthof Approach
取合部施工延長
296 500
200 Gravity
Retaining Wall
40 950
重力式擁壁
600
31 300 31 000
SteelFoundation
鋼 製Spread
直接基礎
Box-girder Bridgewith Steel Deck
鋼製支持梁
鋼床版箱桁
ﾃｰﾙｱﾙﾒTerreSteel 鋼Spread製 直Foundation
接基礎
Armee
交差点
Intersection(on CL)
Length of Harada
Viaduct Bridge
原田高架橋
橋長 192 500 (CL上)
Girder Length
桁長 192 100 (CL上)(on CL)
16 000 18 000 16 000
46 000 (CL上)
21 000(on CL)

1200
4431

取合部施工延長
Length of Approach
371 000
Gravity Retaining 200

防護コン

28500

Improvement

P5

P6

A2
（鉄

Fig. 9 Construction Plan for the Harada Viaduct
Table 1 Comparison of Construction and Traffic Restrictions Period
Month

1

2

3

4

5

Work Period Conventional methodforEntire Project UFO Method(¥100 million)

7

8

9

10

11

12

13

14

15

16

17

18

19

20

21

22

23

450 Days (15 Months)

Saving of 300 Days (40% Reduction)

630 Days (21 Months)
Saving of 380 Days (60% Reduction)

250 Days (8.3 Months)
315 Days (10.5 Months)
185 Days (6.2 Months)

Saving of 130 Days (40% Reduction)

Difference in Losses Due toEnvironmental Degradation, Delays in
Goods Distribution, etc.

Difference in Losses during Work Period

20.0

Difference in Losses during Traffic Restriction Period

+α

15.0

1.5
1.2

10.0

13.7
5.0

24

750 Days (25 Months)

Work Period Conventional methodforOverpass
UFO Method
Overpass Work Conventional method
Traffic
Restrictions
UFO Method

6

13.4

11.0

13.4

Construction
Cost

0.0
Main Work (Conventional)

UFO Method

Fig. 10 Economic Benefits of the UFO Method (Evaluated by MLIT)

158

25

26

(on CL)
Length of Overpass
Section 561118 (CL上)立体区間長
Length of Bridge Section(Bridge Length)(on CL)
橋梁区間長（橋長)142000(CL上)Lengthof Approach 214813
取合部施工区間長

34838
Gravity
重力式擁壁

110000T-shaped
Ｔ型擁壁

150

Retaining
Wall

Retaining
Wall

600

Lengthof Approach 204305
取合部施工区間長(on CL)
Girder Length
桁長141700(CL上)(on CL)
62000 (CL上)

39250

39250

Box-girder
Bridge with Steel Deck
鋼床版箱桁

150

100000T-shaped
Ｔ型擁壁

34838
Gravity
重力式擁壁

600

Retaining
Wall

Retaining
Wall

Steel
Pier
鋼製橋脚
Steel Spread
Foundation
鋼製直接基礎

A1

7500

7500

P1

P2

Steel
Spread Foundation
鋼製直接基礎A2

Steel Girder
Steel Pier

Steel Spread Foundation(Isolated Footing)
RC SlabProtection ConcreteFig. 11 Structural Overview of the Tonyamachi Overpass

Main Girder
Foundation
Steel Pier
Alluvial Gravel Layer(N value: 30 to 50)
Alluvial Gravel
Layer(N value: about 30)
Cohesive Soil Layer(N value: about 10)

Fill Layer

Ground: Solid Element

Fig. 12 Three-dimensional FE Model (Harada Viaduct)

Main Girder
Supporting Beam

Steel Pier

Underneath
Footing

Evenly Dispersed at Depth ofApprox. 6 to 7 m

RC SlabFig. 13 Example of Analysis Result (Subgrade Reaction Distribution)

159

Steel Pier

Steel Spread Foundation

GroundSteel Pier

Steel Spread Foundation(Supporting Beam)

Steel Spread Foundation(Connecting Beam)

1500
1500

42557

11000
8500

82600
79700

Pelast. : Subgrade Reaction Intensity of Steel Spread Foundation as an
Elastic Body
Prigid : Subgrade Reaction Intensity of the Footing as a Rigid Body

PrigidPmax
rigidelast.− Pmean
≤ 20
%
±20%
Pmean
rigid

Pelast.
PrigidP=

Dispersivity of Subgrade Reaction

Pelast.

Fig. 14 Three-dimensional FE Model (Tonyamachi Overpass)

Direction perpendicular
to bridge axis

Bridge axis direction

Fig. 15 Judgment of Rigidity Body

Coordinates of Direction
Perpendicular to Bridge Axis (m)
橋軸直角方向座標（m）

G1

G2

Reaction of Steel Spread各要素の地盤反力度
Foundation
200
0
35
-200
-400

37

PPmeanrigid

Subgrade
Reaction (kN/m2)
地盤反力度(kN/㎡)

40011.8 mDirection Perpendicular to Bridge Axis

Level 2 Earthquake

39

41

43

45

0.8×P
0.8×Prigid
mean

1.2×P
rigid
1.2×Pmean
-600
-800
-1000

Reaction dispersivity when footing is剛体とした場合の反力分散度 Pmean
rigid body Prigid
Steel spread foundation

-1200

8.9 m

Bridge Axis(b) Subgrade Reaction Distribution(a) Subgrade Reaction Contour

Fig. 16 Example Judgment Result for Footing Rigidity

160

47

25th US-Japan
Bridge Engineering Workshop

Session4

Remedial Work & Partial Replacement
Experimental Study on the Time Dependent Flexural Behavior of Prestressed
Reinforced Concrete Beams
By Hiroshi Watanabe, Hirohisa Koga, Hisashi Aoyama, and Yuuki Takeuchi
Overnight Delivery - NJDOT Rapid Bridge Replacement
By X. Hannah Cheng and Harry A. Capers, Jr
Effect of Reducing Strains by SFRC Pavement on Ohira Viaduct
By Takayoshi Kodama, Mamoru kagata, Shigeo Higashi, Kiyoshi Itoh, andYatsuhiro Ichinose
Rapid Bridge Repair / Rehabilitation in Washington State
By Jugesh Kapur

Page Left Blank

Experimental study on the time dependent flexural behavior of prestressedreinforced concrete beams
H.Watanabe1, H.Koga1, H.Aoyama2, Y.Takeuchi2
ABSTRACT
For prestressed reinforced concrete structures, cracking is allowed at theserviceability limit, and the crack width should be controlled by deformed reinforcing
bars and prestressing forces. There are many research results relating to the flexural
crack widths in RC beams; however it is not clear in PRC beams especially the effectof creep and drying shrinkage on their crack width. Experimental study using PRC, RC
and PC beams are carried out to investigate it. In this paper deformation and crack
widths of PRC beams under the sustained flexural load will be discussed with the dataof creep and drying shrinkage of concrete without reinforcement.
INTRODUCTION
In prestressed reinforced concrete bridges, cracking is allowed at theserviceability limit and the crack width should be controlled by arrangement of
de-formed reinforcing bars and prestressing forces. PRC bridges increases in number
recently while most prestressed concrete road bridges in Japan were designed to avoid
cracking in design load combinations.
In PRC bridges, crack control is important to as-sure durability. Howeverlong-term behavior of crack width in PRC beams, especially the effect of creep anddrying shrinkage, is not clear.
In this paper, cracking, deformation and crack widths of PRC beams under thesustained flexural load will be discussed with the data of creep and drying shrinkage ofconcrete without reinforcement.
EXPERIMENTAL PROCEDURE
Dimensions of test beams are shown in Figure 1 and Table 1. Span length ofeach beam is 3000mm.
Beam A1 was designed as fully prestressed concrete member; there is nocompressive/tensile stress at the concrete surface with deign load in this test, 27.5kN
m (1.0Md). Beam B1, B2 and C2 were designed as PRC members and prestressing
force of these beams were decreased. Beam D1 and D2 were designed as RC members.
Diameter of re-bars was selected as tensile stress of longitudinal reinforcements
in beam B2, C2 and D2 would be approximately 200MPa with 2.0Md bending
1

Senior Researcher, Center for Advanced Engineering Structural Assessment andResearch, PWRI
2Former Interchange Researcher, Advanced Engineering Structural Assessment andResearch, PWRI

163

Table 2. Properties of prestressing bars and rebars.
Beam Specimen Diameter
E Modulus
Yield Strength<Prestressing bars>
A1
17mm
200GPa1061MPa
B1, B213mm
201GPa1055MPa
C2
9.2mm
200GPa1262MPa<Re-bars>
A1
10mm
187GPa369MPaB1, B216mm
188GPa365MPaC2
19mm
185GPa379MPaD1, D222mm
186GPa386MPaTable 3. Properties of concrete
Water cement ratio
49%
Compressive strength
39.7MPa
E Modulus
28.6GPa
Tensile strength
3.22MPa164

4555 155 45

300

300
moment.
Prestressing tendons and
Prestressing bars
steel reinforcing bars were incompliance with JIS. Prestressing
force was introduced after 10 days
25
25
from casting. Properties ofprestressing tendons, tensile
longitudinal reinforcement andconcrete are shown in Table 2 andRe-bars
Table 3.
50100 10050
Sustained load test started
Figure 1. Section of test beams.
approximately 28days after casting.
Test beams were set upside Table 1. Test beams.
___________________________________________________
down on support steel
Beam
A1
B1
−
D1
H-beams and flexural load
B2
C2
D2___________________________________________________
was introduced by four
Prestressing bars* 2-17mm 2-13mm 2-9.2mm −prestressing tendons placed Re-bars*
2-10mm 3-16mm 3-19mm 3-22mm
Prestressing force 272kN
159kN
80kN
−
near both ends of beams
k**
100%
58%
29%
0%
(Fig. 2). Sustained flexural
Sustained load***1.0Md (A1, B1 and D1)
load was checked and2.0Md (B2, C2 and D2)
adjusted at least once a
Tension stress 1.6
38.0
−
110.8
of Re-bars (MPa) −
222.0
219.0
210.8
month.
Tensile stress 0.0
2.1
−
4.8
Displacement at the of concrete
−
7.5
8.7
9.8
mid-span of test beams,
surface(MPa)___________________________________________________
strain of prestressing steel,* Diameter of prestressing bars/re-bars
re-bars and concrete were** k = M0 / Md, M0 is the bending moment with which there isnostress at concrete surface in tension side.
monitored with strain*** 1.0Md = 27.5kN m in this test. Sustained loads in beam B2,
gauges. Deformation of
C2 and D2 were adjusted to control maximum tensile stress inbeams was measured withtensile longitudinal reinforcement (200MPa).
contact strain gauge that150

200 300

1350

test beamsteel H-beam

300

300 200

3000

prestressing bar（13mm in diameter）

Figure 2. Arrangement for sustained load test (dimensions in mm)：for contact strain gauge：strain gauge on re-bars：strain gauge on concrete

300<top view>

displacemen gauge<front view>
950

9@100=900

950

300

900

100
35 145
10 65 45

δ

100

1200

900

Figure 3. Positions of strain gauges and points for measurement (dimensions in mm).
Table 4. Drying shrinkage of concretehas 100mm base length
Days after castingand in cracked area
28
62
91
227
413
measured values are taken
Strain × 10-6
97
156
216
348
402
as the change of crack
Drying started 4days after casting.
width. This measurement
was carried out once a
Table 5. Creep coefficientmonth. Positions of strain
Days after introducing
prestressing force 22
64
201
386
gauges and points forcoefficient
0.73
1.04
1.60
1.84
measurement are shown in
Figure 3.
Drying shrinkage of concrete was measured at center of 300 × 300 × 1200mm
concrete specimen (Table 4). Creep coefficients were measured with 300 × 300 ×3600mm concrete specimens with 483kN prestressing force (Table 5).
TEST RESULTS
Cracking of beams
Cracking maps after one year loading are shown in Figure 4. Nine or ten cracks
were observed on the surface of beam B2, C2, D1 and D2 at the start of sustained load
test. The numbers of cracks have not changed through the test. Five cracks wereobserved on the surface of beam B1 after 60days loading while no crack was observed165

at the beginning of test. The
number of cracks in beam B1
has not changed after cracking
was observed.
In beam B2, C2, D1 andD2, average crack spacing ineach beam is fit to thecalculation result pro-posed in
JSCE standard specificationthat is based on an assumptionthat tensile stress in concrete is
negligible. On the other hand,in beam B1, number of crack isless than that of B2, C2 and D2.
No crack was observedon the surface of beam A1.

216 221 173

<B1>

271

*)a
74

<B2>

166

*)b

115 130 166 11976
109 102*)b*)a

129 151 189 90
196
146
154
83

<C2>
30

*)b
<D1>

*)a

173 224 152 258 193 118

*)b
<D2>

127 199 126
141

*)a

*)a
183 114
73 158*)b*)a:Max crack width of concrete*)b:Max strain in re-bar

Figure 4. Cracking maps after one sustained load test.
A1
B1
D1

B2
C2
D2

2

deflection (mm)Deflection of beams
Mid-span deflections oftest beams are shown in Figure
5. Deflection of beam B2, C2
and D2 has been increased
approximately 0.5mm undersustained flexural load anddeflection of beam D1
in-creases approximately
0.4mm.
Beam A1 and B1 haveno bending crack at the
beginning of sustained load test
and deflections of these beams
are almost same. However,deflection of beam B1 has
increased through the sustained
load test while deflection ofbeam A1 has not been
in-creased significantly.

1.5

1

0.5

Crack width0
Maximum crack widths
0 50 100 150 200 250 300 350 400
of test beams are shown indays after start of sustained load test
Figure 6. Crack widths
Figure 5 Change of the mid-span deflection of eachincreased in first 6 months andbeam.
then show some decreasing.
The reason of this is not clear,but temperature change can affect. Maximum crack widths of beam B2, C2 and D2
loaded to have almost the same tensile stress in tensile longitudinal reinforcement have

166

B1
D1
B2

C2
D2

crack width (mm)

0.4

0.3

0.2

0.1
cracking was observewd in B10050 100 150 200 250 300 350 400
days after start of sustained load test

Figure 6. Change of the maximum crack width of eachbeam.
period of sustained load test
1day
21days200days386days

0.4
0.3
crack width (mm)been increased approximately
0.1mm through sustained
flexural test. The effect ofprestressing force to thelong-term behavior of crack
width is not clear.
Maximum crack widthsof beam D1 on which the half ofD2 bending moment wasapplied has been increased
approximately 0.07mm.
While there is no crack
in beam B1 when the sustained
load test was started, changing
of measured lengths with thecontact strain gauge arecalculated and maximum crack
width has been increased
approximately 0.04mm.
Measured crack widths
in beam B2 are shown in Figure
7. Increased crack widths insustained load test have beendifferent in each measured point.
However, there are relatively
bigger cracks and smaller cracks
on the surface of beam B2 andthis tendency has been kept
through test period.

0.2
0.1
0
-0.1

Strain of concrete and tensile
cracking-0.2
longitudinal reinforcement
-500
0
500
distance from the center of the beam
Strain at concretesurface in four sections andFigure 7. Change of the distance (crack width) on thetension side concrete surface of beam B2.
strain of tensile longitudinal
reinforcement in ten sections aremonitored through the test. In
Figure 8, average values of these data in beam B1 and B2. In Figure 8, results of linear
regression analysis with strain of concrete are shown.
In beam B2, compressive strain of concrete has increased approximately 900 ×-6
10 and strain of tensile longitudinal reinforcement has increased approximately 500
× 10-6 through sustained load test. Change of strain of tensile longitudinal
reinforcement corresponds with change of regression line calculated with strain of
concrete. Also in beam C2, D1 and D2, strain of section looks to be proportional to the
distance from the neutral axis.
In beam B1, when sustained load test started, strain of tensile longitudinal
reinforcement is almost zero as there was no crack in beam B1 and tension force was167beam B1

300
250
200
150
100

1000

500

0

-500

-1000

0

-1500

50
-2000

distance from the compression
side of beam (mm)180days386daysbeam B2

300
250
200
150
100

1500

1000

500

0

-500

-1000

0

-1500

50
-2000

distance from the compression
side of beam (mm)strain of concrete/passive reinforcement

strain of concrete/passive reinforcement

Figure 8. Change of the distance (crack width) on thetension side concrete surface of beam B2.
B1
D1
B2

C2
D2

15

-6

curvature (x 10 /mm)

Distribution of strain in
tensile longitudinal
reinforcement
From thedifference of numbers ofcrack, crack widths andstrain in tensile
longitudinal
reinforcements between
beam B1 and B2, effect oftension stiffening would
be different in two beams.
To discuss it, distribution
and change of strain oftensile longitudinal
reinforcement where equalbending moment is loaded
are shown in Figure 10.
In beam B1, strainof tensile longitudinal
reinforcement was
uniform when sustained
load test was started andthis uniformity was kept
before cracking while
strain of reinforcement

1day
21days1500

mainly sustained by
concrete. Strain of tensile
longitudinal reinforcement
in beam B1, however, hasbeen increased with time.
This should be the effectof cracking and decreaseof tension force shared by
concrete in tension area.
From the data ofstrain of concrete andtensile longitudinal
reinforcement, deflectionof each beam wascalculated as curvatureand shown in Figure 9.
Change of curvature is ingood accordance with
mid-span deflection
shown in Figure 5.

10

5

0
0

50 100 150 200 250 300 350 400
days after start of sustained load test

Figure 9. Change of mid-span curvature calculated with thedata of strain gauges on concrete and tensile longitudinal
reinforcement.

168period of sustained load test
1day
21days200days386daysstrain of passive

reinforcement (x 10-6)

1000
800

cracking

600
400
200
0
-500

0
500
distance from the center of the beamstrain of passive-6

reinforcement (x 10 )

2000
1500
1000
500
0
-500

cracking
0500
distance from the center of the beam

Figure 10. Distribution of strain of tensile longitudinal
reinforcement.
50
0-50
strain x 10-6

increased with time. After
cracking, strain of
reinforcement has increased
more in the area near bending
crack.
Cracks in B1 wereobserved after 60days loading
and the tensile creep canaffect the cracking. Tensile
strength of concrete is known
to become smaller whenloading rate is small. In B1,
tensile stress of concrete
under the sustained flexuralload is closet to the tensile
strength of concrete.
In beam B2, there isno newly cracking under thesustained load and change ofstrain of tensile longitudinal
reinforcement under thesustained load is al-mostsame in each strain gauge. Inbeam C2 and D2 under 2.0Md
sustained load, also in beam
D1 under 1.0Md sustainedload, strain of tensile
longitudinal reinforcementchanged alike. Bond between
concrete and tensile
longitudinal reinforcement
has not changed in thesebeams.

-100

distance from the concrete surface(compression side )
110mm
40mm
10mm

-150

-200
The effect of drying shrinkage-250
of concrete
One of the beam
-300
specimen, arrangement of-350
reinforcements is the same as
0
50 100 150 200 250 300 350 400
days after strat of sustained load test
B1 and B2, has been placed
upside-down on steel H-beam
Figure 11. Strain of concrete in the test beam that is notwithout prestressing force
loaded.
and sustained load. Drying
shrink-age of this test beam is shown in Figure 11. Progress of drying shrinkage inthree monitored point of this test beam is similar to that of plain concrete shown in
Table 3.
Drying shrinkage of plain concrete after through the period of sustained load169

test is 305 × 10-6 (Table 3) and maximum space between cracks in beam B2, C2 andD2 is 199mm (Fig. 4). Increase of crack width caused by drying shrinkage is estimatedto be approximately 0.06mm. However, this calculation result is half of actual crack
widening in beams under 2.0Md sustained load (Fig. 6). Crack widening undersustained load test, therefore, can not be explained by drying shrinkage only.
COMPARISON WITH PROPOSED ESTIMATION METHOD FOR CRACKWIDTHS
In JSCE (Japan Society of Civil Engineer) standard specification for concrete
structure, examination for flexural crack width should be done with Equation 1 andgiven permissible crack width. In other specifications for PRC bridges, framework of
examination method is similar.
⎛σ⎞w = 1.1k1k 2 k 3 {4c + 0.7(c s − φ }⎜⎜ se + ε ' csd ⎟⎟
⎝ Es
⎠

(1)

where, k1,k2,k3 = constant values to take into account the effect of surface geometry of
reinforcement, concrete quality and multiple layers of tensile reinforcement on crack
width; c = concrete cover(mm); cs = center to center distance of tensile reinforcement(mm); φ = diameter of tensile reinforcement (mm); ε'csd = compressive strain forevaluation of increment of crack width due to shrinkage and creep of concrete; σse =increment of stress of reinforcement from the state in which concrete stress at the
portion of reinforcement is zero N/mm2.
However, the value of ε'csd for designing PRC bridge has not been establishedyet. To compare with proposed design values as ε'csd, ε'csd in each beam is calculatedwith maximum crack width and increase of tensile longitudinal reinforcement in eachbeam and shown in Figure 12, as calculated value of w in Equation 1 virtually shows
maximum crack width.
Calculated values of ε'csd after one year sustained load test range are bigger thanproposed value, 150 × 10-6 by JSCE in beam C2, D1 and D2. Calculated values of ε'csd
is smaller in PRC beam than RC beam. However the cause of this tendency is not clear.
B1

150

ε'csd

beam

D1
B2

εs

289

240

627

125

C2

1229

208

D2

1384

240

0

1425

500

1000 -6
strain x 10

1500

2000

Figure 12. Calculated values of ε'csd and εs after 390 days sustained load test.

170

CONCLUSIONS
Increase of the mid-span deflection and bending crack width are observed inbeams under sustained flexural load and prestressing force dose not affect the rate of it
when tensile strain of tensile longitudinal reinforcement is same.
Strain of concrete and tensile longitudinal reinforcement in beams undersustained flexural load has been proportional to the distance from the neutral axis
generally.
Bond between concrete and tensile longitudinal reinforcement has not been
affected significantly by sustained flexural load.
Compressive strain for evaluation of increment of crack width due to shrinkage andcreep of concrete (ε'csd) was bigger than proposed value by JSCE in some test beams.

References
JSCE (2002) Standard specifications for concrete structures-2002 “Structural
Performance Verification”

171blankOVERNIGHT DELIVERY –NEW JERSEY DOT RAPID BRIDGE REPLACEMENT
Xiaohua H. Cheng1 and Harry A. Capers, Jr.2
Abstract
The aging highway bridge infrastructure in the US is being subjected toincreasing traffic volumes, structural deficiency and functional obsolete. They must becontinuously renewed while accommodating traffic flow. This paper demonstrates
application of prefabrication technology where traffic control issues demanded rapid
construction and decision making risk of total facility showdown. It shows an example
supporting the FHWA’s Accelerated Bridge Construction Decision Making
Framework. Using the lessons learned from previous experience with the emergency
repair/replacement, the Rt.1 reconstruction project involved replacing superstructuresof two bridges with prefabricated components installed in 59-hour time window over aweekend. The experience becomes standard practice of “Hyperbuild”.
Introduction
The aging highway bridge infrastructure in the United States is being subjected
to increasing traffic volumes, structural deficiency and functional obsolete. They must
be continuously renewed while accommodating traffic flow. The traveling public isdemanding that the rehabilitation and replacement be done more quickly to reduce
congestion and improve safety. Conventional bridge reconstruction is typically on thecritical path because of the sequential labor-intensive processes of completing thefoundation, the substructure, the superstructure components, railings, and otheraccessories. Prefabricated bridge systems can allow components to be fabricatedoff-site and moved into place quickly while maintaining traffic flow. Depending on thespecific site conditions, the use of prefabricated bridge systems can minimize traffic
disruption, improve work-zone safety, minimize impact to the environment, improve
constructability, increase quality, and lower life-cycle costs. New Jersey Department ofTransportation (NJDOT) clearly demonstrated the truth of these statements with theirapproach to replacement of the superstructures of two structurally deficient bridges
carrying a freeway section of US Rt.1 through the Capital City of Trenton.
The project involved completely replacing three superstructures of two bridges
with new ones designed for a 75 year life, made off site and installed over three
weekend shutdowns of 59 hours each. The design and construction of this project
learned lessons from the past experience of two emergency repair/replacement projects.
A project of this magnitude would typically take the Department approximately 2 years
to design and construct as a traditional deck replacement and bridge rehabilitation
1

Senior Engineer, Bureau of Structural Engineering, New Jersey Department ofTransportation (NJDOT), Trenton, NJ 08625, USA
2
Vice President, Arora and Associates, P.C., Lawrenceville, NJ 08648, USA173

project. The project approach saved more than 22 months and an estimated design andconstruction cost, including delay-related user costs in excess of $2M. The results? An
extremely happy motoring public, bridge owner and contractor.
This paper will outline the application of prefab technology on the project wheretraffic control issues demanded rapid bridge construction techniques. The paper will
also demonstrate a past project where decision makers took the risk of total facility
shutdown to allow for rapid emergency repair/replacement as lessons learned for thisproject.
Needs for Accelerated Bridge Construction
Bridge engineers have successfully used accelerated bridge construction (ABC)practices for many years around the globe. Our good fortune in this country toexperience continued growth has also had the result of developing a greater dependence
on our transportation infrastructure and less tolerance to interruptions caused by taking
lanes out of service for routine maintenance. With the advent of high performancematerials and emerging advanced technologies, the FHWA is attempting to provide
leadership in meeting the public’s expectations as illustrated in their Vision andMission Statements. They identified three “Vital Few” important items to be focusedon by the agency, those being Safety, Congestion Mitigation, and Environmental
Stewardship and Streamlining.
In focusing on these priorities and goals the FHWA has led to therecommendation for modular prefabricated construction, among other things. The
concept of prefabricated bridge elements and systems (PBES) is being researched aswell as applied and put to use in building bridges as illustrated in the FHWA Decisionmaking Framework (FHWA, 2005)
To obtain information about technologies being used in other industrializedcountries, a scanning tour of five countries was made in April 2004 (Ralls et. al, 2005).
The overall objectives of the scanning tour were to identify international uses ofprefabricated bridge elements and systems and to identify decision processes, design
methodologies, construction techniques, costs, and maintenance and inspection issues
associated with use of the technology. The prefab bridge elements and systems
consisted of foundations, abutments, piers or columns, pier caps, beams or girders, anddecks. Bridges with span lengths in the range of 20 to 140 feet (6 to 43m) were themajor focus, although longer spans were of interest if a large amount of innovative
prefabrication was used.
The focus areas of the study were, therefore, prefabricated bridge systems that
Minimize traffic disruption,
Improve work zone safety,
Minimize environmental impact,
Improve constructability,
Increase quality, andLower life-cycle costs.

174

During the tour, it was observed that precast deck panels were used on steel
beams to produce both composite and non-composite members. The use of full depth
prefabricated concrete decks reduces construction time by eliminating the need toprovide cast-in-place concrete. Composite action was developed through the use ofstuds located in pockets in the concrete deck slab. In addition, use of innovativeequipment, e.g. self-propelled modular transporters (SPMTs) were observed. These
provided a means to accelerate bridge construction using a factory produced product forthe state bridge owners.
According to the Year 2006 National Bridge Inventory (NBI), New Jersey has
nearly 6,500 highway bridges (2,580 NJDOT bridges). About 35% are structurally
deficient and functionally obsolete. The average age is close to 50 years old, while 52.5
years old for steel bridges. The most projects under plan, design, and construction are
rehab and replacement related. Furthermore, New Jersey is the most highly urbanizedand densely populated state in the US. Therefore, rapid bridge construction for both
new and old bridges using prefab elements and systems is extremely vital in New
Jersey.
In 2004 NJDOT established “Hyperbuild” program for accelerated roadway andbridge constructions. The term “Hyperbuild” was coined by the NJDOT Commissioner
Jack Lettiere. Hyperbuild projects acknowledge the tremendous need to minimize
traffic impacts for all of the previously mentioned reasons and recognize the potential
savings of millions of dollars in design, construction, and road user costs that could be
realized. The Commissioner’s vision was to reduce the time from initiation of design tothe opening of the finished project to traffic (“project end date”) and require roads befunctional while constructing, by utilizing innovative methods of design andprocurement for certain types of projects. Such projects have to have a well-defined
scope and, if possible, require limited right-of-way acquisition, utility relocations andenvironmental impacts. With the past experience of rapid emergency replacement of
I-295 project, the US Rt.1 Freeway reconstruction became the first “Hyperbuild”
project. These two projects will be presented in the following sections, respectively.
Rapid Emergency Bridge Replacement
Creek Road Over I-295 Bridge Emergency Repair/Replacement
In May 30, 2002, a tractor trailer traveling along I-295 South struck the overpass
for Creek Road over I-295 in Bellmawr, Camden County, New Jersey near the borderof Gloucester County (NJDOT, 2002a). The location is very sensitive because theestimated ADT in that area of I-295 is 80,000-120,000. The impact of the oversized andover height piece of earth-moving equipment that the tractor-trailer was carrying
shattered the fascia beam and the fifth interior beam causing serious damage andpotential catastrophic failure of the overpass (Figure 1). The overpass was a two
simple span bridge having eight prestressed concrete girders and was classified
structurally deficient (SD) and functionally obsolete (FO) with the permit

175underclearance of 13’-9” (4.19m) per the inspection prior to the accident, while NJDOT
currently requires 16’-6” (5.03m) underclearance for bridge design.
After initial inspection by the NJDOT bridge staff, the overpass was found to be
structurally unsound requiring closure of the both I-295 mainline and Creek Road until
temporary repairs could be made. The bridge was strengthened with a new, reinforced
beam that was installed from the top of the bridge deck (strongback). Options to repair
the bridge from under the deck were evaluated, but ultimately rejected, because theywould have further restricted the bridge’s maximum height safety requirements. The
mainline of Route I-295 was reopened after the installation of a temporary support
beam, but replacement of a portion of the deck and structural members down to theabutments was required before Creek Road could be reopened.

Figure 1 Exposed Prestressing Strands of Damaged Interior and Fascia Girder of Creek
Road Overpass Bridge
Creek Road overpass is a major route supporting commercial and emergency
services for the City of Bellmawr and any traffic diversions on Creek Road or from
I-295 onto surrounding network roads would cause unacceptable abnormally heavy
traffic in those areas. Such conditions constituted an emergency in the region and it
was determined by the Governor’s office that damage to the overpass required
immediate repairs on an emergency basis.

176

The Governor’s declaration of an emergency allowed the department’s
structural engineers to quickly search various proprietary prefabricated bridge systems
to rapidly put the bridge back into service (McGreevy, 2002; NJDOT, 2002b). The
chosen strategy was to advertise for a superstructure system consists of a prefabricatedcomposite concrete deck and stringer unit usually produced for simple spansmanufactured by a supplier.
Several stipulations were that the systems would be designed in accordance withcurrent American Association of State Highway and Transportation Official’s Bridge
Design Standards (AASHTO), would utilize weathering steel, concrete with corrosion
inhibitors (high performance concrete (HPC)) to increase service life and would
provide a shallower superstructure depth than existed prior to the accident.
The permanent replacement was made utilizing “Inverset” deck panels (Figure2). The panels provided are cast inverted in forms suspended from the support girders
and when turned upright the decks are precompressed which should provide forincreased durability. Panels come in lengths up to about 100 feet (30.48m) (86.5’(26.37m) for this bridge) with widths usually in the 8 to 12-foot (2.44 to 3.66m) range.
Stringer spacing of the panels was arranged to permit the construction of new bearing
seats under the existing bridge between existing stringers prior to demolition of thedamaged superstructure. It should be noted that these bearing seats were alsoprefabricated to minimize construction time on site. Spacing the stringers to allow thisconstruction of the bridge bearing seats also resulted in closer spacing allowing for a
9-inch (0.23m) increase in vertical underclearance.
Upon completion of off-site fabrication and delivery on site, required
modifications to the bearings, staged demolition of existing structure, erection of the
new modular panels (four units) was performed on successive weekend night
shutdowns, tremendously minimizing traffic impacts to the interstate I-295 and Creek
Road (Figure 3). Besides replacement of the damaged span, the other existing span
was repaired on concrete spalls and previous minor collision impact.

Figure 2 Prefabricated “Inverset” Deck Panel Figure 3 Night Time Erection of Modular(Modular Deck-Girder Unit)
Deck Panels Over I-295

177

Rt.70 Friendship Creek Bridge Emergency Replacement
The Rt.70 Friendship Creek Bridge was a concrete encased steel girder bridgewith concrete gravity abutments at both ends. On July 12, 2004, 13 inches of rain fell ina 12 hour period in the area that was equivalent to a 1,000 year event storm. The storm
caused that six dams were breached and the bridge collapsed with abutment failure on
July 13, 2004 in the early morning due to scour. The State Police and NJDOT closed theroad and established the detour for local traffic. A temporary bridge was built using
prefabricated steel truss and the road opened to traffic on July 19, 2004 (Figure 4). It
was followed by a new bridge construction including new abutments and newsuperstructures. The replaced superstructure is of precast concrete box beams withasphalt overlay. The replacement substructures is PZ steel sheeting with pipe piles
bearing support and tiebacks. The first half of the replacement bridge completed on
September 9, 2004 and the full bridge opened to traffic in October, 2004 (Figure 4).

Figure 4 Temporary Bridge and Replacement Bridge Using Prefabricated system andComponents

US Rt.1 Freeway Bridge Reconstruction Project
Each day the US Route 1 carries more than 50,000 vehicles through the City ofTrenton, the capital of the State of New Jersey. The US Rt. 1 serves as a vital link toadjacent Pennsylvania and is a heavily traveled land service route for thosecommunities along the Northeast Corridor between New York City and Philadelphia.
Just north of the City of Trenton in Lawrence Township, Rt. 1 divides into two Routes,
Rt.1 Business which carries local traffic to points in the city and Rt. 1 Freeway which
was intended to provide faster access to downtown Trenton, the State Government
Offices, and the crossing of the Delaware River between Trenton of New Jersey andMorrisville of Pennsylvania.
The freeway section configuration of the highway was constructed onembankment and provides for two lanes of traffic in each direction divided by a Jersey
Barrier along its entire length creating significant challenges for maintenance of traffic
for any work performed within its limits. However, three bridge decks on the Route 1

178

Freeway mainline, one at the Olden Avenue Connector and two at Mulberry Street, haddeteriorated to the point where they were in need of constant maintenance. Both bridges(3 superstructures) were single-span simply supported steel girder bridges built in 1953
with skew of 56 degrees and 10 degrees, respectively. Although the steel girders andbearings were repainted in 2000, the inspection report and evaluation revealed andclassified the superstructures as structurally deficient due to poor condition of concrete
decks and bearings (Figure 5; Figure 6). The deteriorations included large concrete
spalls of decks with exposed rebars, patch spalls, fine transverse and map cracks
underneath decks, steel girder and stiffener buckling due to prior collision impact,
undermined concrete pedestals and rusted anchor bolts of bearings.

Figure 5 Deteriorated Decks of the Route 1
Bridge over the Olden Avenue Connector

Figure 6 Deteriorated Pedestal of Bearings

In 2005 the replacement of these three bridge superstructures was undertaken in
one project that was to become the NJDOT’s first “Hyperbuild” project. The threebridges in the project were actually located at two points along the mainline of US Rt.1
separated by approximately a half mile. The Rt. 1 Bridge over the Olden Avenue
Connector is a highly-skewed steel girder bridge with concrete deck (Figure 5). The
35.0-ft (10.67m) wide single-span bridge has a span of 86.8-ft (26.46m) and carries
2-lanes of traffic. The Rt. 1 Bridge over Mulberry Street consists of two parallel bridge
superstructures on a common abutment with a median barrier separating each directionof traffic. The 82.2-ft (25.05m) wide bridge has a 60.0-ft (18.29m) long single-span andcarries 4-lane of traffic over Mulberry Street.
Preliminary review of the sites of these deficient bridges indicated that they met
the criteria of “Hyperbuild”. Several prefabrication options were considered forrehabilitation of these bridges based on the past emergency repair experience andavailability of precast products (capers, 2005; 2006). Using the lessons learned during
the previous emergency repairs of the Creek Road Bridge over I-295, NJDOT’s
structural engineers concluded that a similar approach of using prefabricatedsuperstructure modular panels for the rapid replacement of these bad decks would be aneffective strategy due to the very limiting project site space and significant need for
rapid project development/design and construction. Use of SPMTs for an entire prefab179

bridge superstructure system was not an option due to the limited site space toaccommodate either the prefeb units or the SPMT equipment.
Each superstructure was designed using 5 full-length modular segments ofvarying width, each with two Grade 50W steel girders and a 9-inch (229mm) thickcomposite concrete “Inverset” deck system. The 86.8-ft (26.46m) long bridge span overOlden Avenue utilized W36x182 girders, and the 60-ft (18.29m) long bridge spans over
Mulberry Street utilized W30x99 girders. Segments sizes considered thetransportability and erection restrictions associated with urban nature of the project site.
All new structures were designed for 75 year design life.
The 15 modular segments were designed and fabricated in Schuylerville, New
York, assembled at the plant to verify field tolerances, and trucked to an airport parking
lot near the bridge (Figure 7). The segments were required to be on site 24 hours prior
to the start of demolition of the existing bridge. The contract specified highperformance concrete (HPC) to be used for all concrete on the job.

Figure 7 Prefabricated Modular Segment in Transportation
The construction was planned to occur using only weekends to shut downs
mainline traffic to minimize disruption along the corridor. Each of the 3 superstructures
was allowed a 59-hour window commencing Friday evening (7:00pm) with allactivities off the roadway and both lanes re-opened before the morning rush (6:00am)
on Monday. If this window was exceeded, a Lane Occupancy Charge would beassessed, up to $10,000 per day.
As is typical on NJDOT projects, incentives and disincentives) were included
on this project to encourage the contractor to minimize onsite construction time evenfurther than 59 hours per bridge. For the bridge over the Olden Avenue Connector, an
incentive (or disincentive) of $1,500 per hour was specified if the work was completed
in less (or more) than 59 hours, not to exceed a maximum of $27,000. For eachstructure over Mulberry Street, an incentive (or disincentive) of $2,000 per hour was
specified if the work was completed early (or late), not to exceed $36,000.

180Liquidated damages were also specified. The contractor would be charged$4,200 per day if the bridges weren’t substantially completed by the specifiedcompletion date in the contract, and an additional $900 per day if all work was notcompleted within 3 months following that.
The engineer’s estimate for this project was $3.8M. The low bid of $3.5M was
8% or $297,000 less than the engineer’s estimate. There were 5 bidders on this project.
The second lowest bid was 10% higher than the low bid.
The Rt. 1 Bridge over the Olden Avenue Connector was replaced during aweekend closure in August 2005. The Rt. 1 Southbound Bridge over Mulberry Street
was replaced during a weekend closure in September 2005, followed by the Rt. 1
Northbound Bridge over Mulberry Street during a weekend closure in October 2005.
Design and construction of such bridges would have taken 22 months using
conventional methods.
Olden Avenue Bridge The Route 1 Bridge over the Olden Avenue Connector was
closed at 7 p.m. on a Friday in August 2005, and traffic was rerouted onto a 5-miledetour. The bridge was demolished in place using conventional methods. The existing
abutments were repaired and new bearing seats were constructed. The prefabricatedsuperstructure was then erected during nighttime. The longitudinal joints between
superstructure segments were then sealed, and the expansion joints at the ends of the
span were completed. The cast-in-place (CIP) parapets were connected to the outside
segments with bars in threaded inserts.
Mulberry Street Bridge The Rt. 1 Southbound and Northbound bridges over MulberryStreet were closed at 7 p.m. on a Friday in September and October 2005, respectively,
and traffic was rerouted onto a 5-mile detour for Southbound Mulberry, while off- andon-ramps were used for Northbound Mulberry. The construction methods and timerequired to replace these bridges were similar to the bridge over the Olden Avenue
Connector. Parapets were cast-in-place (CIP) concrete and median barriers wereprecast. The typical construction time frame was as follows.
- Weekend closure: Friday 7:00 pm to Monday 6:00 am- New bearing pedestal installation: prior to and during the project- Demolition and removal of old structure: Friday/Saturday (Figure 8(a))- Bearing and expansion joint replacement: Saturday- Prefab modular bridge unit erection: Saturday night (Figure 8(b); Figure 8(c))- Bridge installation completion: Sunday morning (Figure 8(d))- CIP parapets and prefab median barriers: Sunday- Cleaning and Reopen: Monday early morning

181

(a) Demolition/Removal of Existing Structure(b) Initiation of Erection of Route 1 Southbound Bridge over Mulberry Street(c) Erection of Prefabricated Modular Segment

182

(d) Erection of Final Deck Panel of Route 1 Southbound Bridge over Mulberry Street
Figure 8 Reconstruction of Two US Rt.1 Freeway Bridges
All three bridges were opened in less than the required 59 hours. The bridgeover the Olden Avenue Connector was opened in 56 hours, the bridge over SouthboundMulberry was opened in 51 hours, and the bridge over Northbound Mulberry was
opened in 54.5 hours. With all three bridges opened well before Monday morning rush
hour, the contractor earned an $18,500 incentive.
Summary
Each of the 3 bridge superstructures in the New Jersey DOT’s first“Hyperbuild” project was replaced over a weekend, during a total of 6 days in 3
consecutive months. The replacements were completed in significantly less than the 22months required for conventional design and construction, and they were completed
under budget. The design and construction savings, including delay-related user costs,
are in excess of $2M.
Each bridge is expected to see a 75-100 year service life due to the quality of its
prefabricated superstructure, the use of high performance concrete, and the attention
given to connection details. Conventionally constructed bridges have an averageminimum 50-year life in New Jersey.
There are a large number of such bridges to repair and rehabilitate in New
Jersey. The experience of Rt.1 Freeway reconstruction set a good example for the future
work and became standard practice of “Hyperbuild” projects. The project clearly
demonstrated the benefits that can be reaped by applying accelerated construction
strategies: less construction time, increase work zone safety, less maintenance, more
durability, higher quality, reduce user cost and life-cycle cost, and minimize traffic
disruption and environmental impact. And of course, an extremely happy motoring
public! In the mean time, NJDOT looks forward to more and more innovative andcost-effective prefabricated products that can be selected for the future projects.

183

References
1. Capers, H. (2005) Hyperbuild! Rapid Bridge Construction Techniques in New
Jersey, Transportation Research Board 6th International Bridge Engineering
Conference, Boston, MA, July, 2005
2. Capers, H., Lopez, J., Lewis, R. and Nassif (2003), H., Applying The Principals
Of Get In, Get Out, And Stay Out Designing And Detailing Of Highway
Structures For Rapid Construction In New Jersey, 2003 NYC Bridge
Conference, New York, NY, August, 2003
3. Federal Highway Administration (FHWA) (2005), Framework forPrefabricated Bridge Elements and Systems (PBME) Decision-Making, Dec.,
2005
4. McGreevey, E. James, Governor, Executive Order #28, State Of New Jersey,
Trenton, NJ
5. New Jersey Department of Transportation, News Release, Creek Road Bridge
Repairs Move Forward to Allow Reopening on Monday Morning, Trenton, NJ,
June 5, 2002
6. New Jersey Department of Transportation, News Release, Creek Road Bridge
Over I-295 Damaged in Afternoon Accident, Trenton, NJ, May 30, 2002
7. Ralls, Tang, Bhidé, Brecto, Calvert, Capers, Dorgan, Matsumoto, Napier,Nickas, Russell (2005), Prefabricated Bridge Elements and Systems in Japanand Europe, Washington, D.C., March, 2005
8. Roads and Bridges Magazine, November, Hyperbuild Project, Scranton GilletteCommunications, Inc. Arlington Heights, II, November, 2005

184

Effect of reducing strains by SFRC pavement on Ohira Viaduct
Takayoshi Kodama1, Mamoru kagata, Shigeo Higashi, Kiyoshi Itoh, Yatsuhiro Ichinose

Abstract
Fatigue damage of orthotropic steel bridge decks owing to increasing traffic
volumes and higher wheel loads have been reported. Steel fiber reinforced concrete(SFRC) pavement holds promise as a countermeasure for improving the fatigue
durability of existing orthotropic steel decks. Static loading tests using a 200 kN dump
truck were conducted on the orthotropic steel deck of the Ohira Viaduct. This paper
reports the results of strains measurement before and after construction of the SFRC
pavement, which show a marked effect in terms of reducing local strains in the steel
deck.

Introduction
There have been numerous reports of fatigue-induced damage and cracking
concerning orthotropic steel deck bridges used by heavy vehicles. The reported crack
occurrences were in the welds between transverse ribs and longitudinal ribs, the welds
between longitudinal ribs and deck plates, the welds between neighboring longitudinalribs, etc. Rare instances of cracks that originate in the vicinity of a deck plate and run
through that deck plate, as well as road surface subsidence [1, 2], also have beenreported, raising the possibility of traffic accidents and making this an important issue
to be addressed.
A number of studies have been conducted on how to prevent such fatigue damage,including the use of highly rigid steel fiber reinforced concrete (hereafter, SFRC) forthe pavement of the steel deck to inhibit local deformation of the steel deck, and the
reduction of local stress to improve fatigue durability. SFRC pavement was used a longtime ago by the Nagoya Expressway Public Corporation [3] as a way to prevent
pavement damage of ramps with a steep longitudinal gradient, tollgates, etc. Morerecently, as a fatigue countermeasure for steel decks, SFRC pavement was applied
through a new method using epoxy resin adhesive instead of conventional stud dowels
for the joints between the SFRC and deck plate, in the section of National Highway 357
forming the lower level of Yokohama Bay Bridge [4, 5], Shonan-Ohashi Bridge [6, 7],
and the Metropolitan Expressway [8].
The Ohira Viaduct, an orthotropic steel deck bridge currently in service in Oyama,
Tochigi Prefecture, features SFRC pavement as part of various fatigue
countermeasures. In the present study, we used this location to run truck-loading tests
on old-type asphalt pavement and new-type SFRC pavement as shown in Fig.1 andFig.2. These tests revealed the effectiveness of SFRC pavement in reducing stress on
the steel deck, as reported below.

1

Manager, Technology Department, KAJIMA ROAD.,LTD185

Influence of an overlay materialFirst of all, the result of researching the adhesion of an adhesive and fresh concreteis shown below.
(1) Three type of overlay materials used for the experiment
It was compared with the adhesion from the difference of overlay materials (cement
paste, mortar and concrete) to hardening base concrete. The used mixture proportionsof base concrete are shown in Table 1. The used cement paste as overlay material was
made into 47% that is the same water cement ratio as the concrete shown in Table1.
The used cement mortar as overlay material was used as water cement ratio 0.47 of
cement paste, and three kinds of sand-cement ratio such as 0.1,0.5, and 2.39 wereadded. Furthermore, overlay concrete that had sand-total aggregate ratio from 40% to
49% was also used. The examination was executed after 7days from producing the test
cores. 0.7kg per 1square-meter of epoxy-adhesives were applied. The time intervalfrom spraying the adhesives to overlaying the fresh concrete was made into 1 hour sothat it might achieve the same conditions as real construction.
Table 1 Mixture proportions of concrete
Gmax
(mm)

W/C
(%)

s/a
(%)

Ad
Cx%

Unit contents(kg/m3)
W

C

S

G

20
47.0 44.4
157
334
797
1030
0.7
Gmax means the maximum size of aggregates. Small size figure of s means the volume of sands anda means the volume of aggregates. Cx% means the content percentage of admixture to cements.

The test result of bonding strength by different overlay materials is shown in Fig.1.
The bonding strength of cement paste and the mortar of 0.1 and 0.5 is a small value, andthey had exfoliated in the bonding interface. With the mortar of 2.39 and concrete, it
had excellent bonding strength and the position of rupture was mostly inside the mortar
or concrete. From the result of this bonding test, it is thought that existence of
aggregate is indispensable in bonding with adhesives and fresh concrete. Aggregatesbonded with adhesives first, paste up firmly, and it is thought that concrete hardens andunifies after that. Fig. 2 shows the bonding strength in the case of sand-total aggregate
ratio in the range between 40% and 49%. In this range, bonding strength was almost thesame. Therefore, it is thought that there is almost no influence that the rate of fineaggregate has to bonding strength.
4

2.5

2

Bonding strength (N/mm )2

Bonding strength (N/mm )

3.0

2.0
1.5
1.0
0.5
0.0
Cementpaste

MortarMortarMortar Concrete
S/C=0.1 S/C=0.5 S/C=2.39

3

2

1

0
37

40

43
46
Sand percentage (%)

49

Fig. 2 Relationship of sandpercentage and bonding

Fig. 1 Bonding strength ofcement paste and mortar

186

52

5

6

2

4
3
2
1
0

2

7

Compressive strength (N/mm )6
Shear strength (N/mm )2

Bending strength (N/mm )(2)Various intensity of the BONDED concrete by EPOXY ADHESIVE AGENT
Next, the strength property in the case of putting it with the adhesives is shown.
The used mixture proportion of concrete is shown in Table 1.The cylinder specimen
with diameter of 100mm and height of 200mm was used for the tests of compressive
strength and shear strength. The height of base concrete and overlay concrete was
100mm in 200mm high. The speed of loading in shear test was controlled bydisplacement to be set to 1mm per 1 minute. The specimen used for the bending test
was 100mm in height, 100mm in width, and 400mm in length, and the base concrete
was set to the lower layer. 0.7kg per 1square-meter was sprayed on base concrete, andfresh concrete was placed 5 hours after the application. The examination was executed
after 14 days from placing the concrete.
The results of strength test are shown in Fig.3. The bonding strength of bonded
specimen was equivalent to a body of base concrete and overlay concrete. Moreover,there was no exfoliation at interface between base concrete and overlay concrete. Shearstrength of bonded specimen is equivalent to base concrete and overlay concrete, andthe rupture position is inside of the overlay concrete. It is thought that this wasdestroyed in concrete because interfacial bonding strengths of the adhesive and theoverlay concrete are larger than tensile strength of overlay concrete. Sufficientadhesion to shear stress was checked. It was shown that it is equivalent to a body ofbase concrete and overlay concrete in compressive strength, and the adhesion byadhesives is absolutely excellent.

5
4
3
2
1
0

Existing Overlay Adhesive
concrete concrete concrete
specimen specimen specimen

Existing Overlay Adhesive
concrete concrete concrete
specimen specimen specimen

60
50
40
30
20
10
0
Existing Overlay Adhesive
concrete concrete concrete
specimen specimen specimen

Fig. 3 Strengths of adhesive bonded specimen
Outline of Ohira Viaduct and SFRC paving
The Ohira Viaduct is a road bridge on National Highway 50 consisting of 31 spans
and measuring 987 m. The bridge's two inbound (Takasaki-bound) lanes werecompleted in 1983, and its two outbound lanes, which run parallel with the inboundlanes, were opened to traffic in 2005. The bridge has independent superstructures forthe inbound lanes and outbound lanes, but common piers support these. The Ohira
Viaduct was paved with SFRC over a length of 213 m, consisting of one 3-spancontinuous 2-box girder and two simple plate box girders of the steel deck with U-ribsof the inbound lanes. The part of the bridge that was selected for measurement in thepresent study consisted of a 3-span continuous steel deck 2-box girder spanning piers187

3 to 6 of the inbound lanes (Takasaki-bound). The deck plates are 12 mm thick, theU-ribs are 320 mm wide, 200 mm high, 6 mm thick, and have an effective span of 40
m, 56 m, and 50 m, respectively. The bridge has been in service for longer than 20
years and the steel deck welds directly under the path of the truck wheels, for example
the welds between the deck plates and U-ribs show signs of fatigue damage. As part of
the fatigue countermeasures, it was decided to use SFRC paving to reduce local stressin the various parts of the steel deck. The SFRC pavement application was split in twoalong the width direction. Stud dowels with a diameter of 9 mm and height of 30 mm
were placed at 300 mm intervals in double rows along the length of the work area (inparallel with the bridge axis) and along the work area joints (at right angles with thebridge axis). The adhesive used was high-durability epoxy resin adhesive that was
shown not to lose bond strength under water-immersed wheel load fatigue testing [9],applied at 1.4 kg/m2 (Photo.1). Further, the adhesive was used not only for the bond
between the deck plate and SFRC, but also as the bond for the joint faces of
SFRC-SFRC construction joints and the SFRC-expansion joint interface.
As the two inbound lanes could be closed to traffic for a period of approximately 1.5
months only, high-early-strength concrete was used for the SFRC. Concrete withdesign compressive strength of 29.4 N/mm2, 20 kg/m3 of expansive agent, and 120
kg/m3 of 30 mm steel fibers were used. To ensure stiffness after cracking undernegative bending moment in the SFRC paving directly above the web of the main box
girder, a 100 mm mesh CRFP (carbon fiber) grid with a width of 1.0 m was placed at
the center of the main girder web at a depth giving a bottom cover of 30 mm (Photo.2).
The CFRP grid that was used had a cross section of 39.2 mm2 and tensile strength of
1400 N/mm2.
To Takasaki

Bridge lengthGauge
attachment

Gauge
attachment

Fig.4 shows a top plan viewFig.5 a cross-section of the bridge section selected for measurement.

188

Photo.1 Adhesive and Stud Dowels

Photo.2 CFRP Grid Placement

Photo.3 Applying epoxy
adhesivesPhoto.4 SFRC Placement

Structural Considerations of SFRC
The main aspects considered are SFRC and the width of the CFRP grid wereconsidered outlined below. Regarding the thickness of the SFRC, three cases wereconsidered, namely the same thickness as the asphalt pavement (75 mm), greater
thickness than the asphalt pavement (100 mm), and a two-layer pavement (SFRC
thickness of 45 mm + surface asphalt thickness of 30 mm), conducting comparisons interms of economic efficiency, workability, and maintainability. In the end, SFRC withthe same thickness of 75 mm as the existing asphalt pavement was selected.
Regarding the CRFP grids and stud dowels for reinforcing the SFRC, two models, theSFRC directly over main girder model and the one bridge portion model were
examined using finite-element analysis. The degree of stress generated in SFRC varies
according to the steel deck structure, including main girder placement, transverse ribinterval, and U-rib interval, and the degree of stress generated by live load, heat of
cement hydration, etc., was obtained with the bridge part model modeling the Ohira
Viaduct.
The effectiveness of CFRP grids for limiting the crack width in SFRC wasconfirmed as a result, and 1.0 m above the main girder was selected as the installation
range. With regard to warping of the pavement owing to heat of cement hydration,
drying shrinkage, etc., adhesive was judged to be a sufficient countermeasure and theuse of stud dowels was deemed unnecessary. However, as the long-term durability ofadhesive in the actual bridge was not ascertained, it was decided to use stud dowels
only at the edges of the SFRC to prevent the infiltration of water, etc., in case of peeling189

off of the adhesive layer. Further, with regard to the future maintenance and repair of
the SFRC pavement, the height of the stud dowels was set to 30 mm (special-orderproduct) and the bottom cover of the grid to 30 mm, in order to support a pavement
cutting depth of approximately 30 mm.
Outline of Fatigue Damage of Ohira Viaduct
In 2005 and 2006, the steel deckpart using U-ribs was inspected andrepaired. As a result, fatigue damage
was found to have occurred innumerous locations, including thefillet welded joints between the deckplate directly under the truck wheels
and the U-rib (Type A), and the fillet
welded joints between the deck plateand vertical stiffeners (Type C).
Figure.6 shows a schematic diagramof the different types of fatigue
damage, and Table 2 lists the number
Type A: Fillet welded joint of deck plate and U-rib
of fatigue damage locations in eachType B: Stud weld of U-rib
Type C: Fillet welded joint of deck plate and vertical stiffener
span. An example of fatigue damage
Type D: Fillet welded joint of transverse rib and U-rib
Type E: Fillet welded joint of deck plate and transverse ribin the Ohira Viaduct is also shown inType F: Fillet welded joint of cross beam/transverse rib flange andPhoto.5. As can be seen in thisphotograph, the crack in the weld Fig.6 Schematic Diagram of Types of Fatigue
between the deck plate and the U-rib Damagechanges direction and progresses into Table 2 Fatigue Damage Locations in Each Spanthe U-rib, and a stop hole is provided
at the tip of the crack.

SectionSuperstructure type

P3-P4
P4-P5

3-spancontinuous
box girder

P5-P6
P10-P1
1
P13-P1
4

Simple steel
plate girder
Simple steel
plate girder
Total

TypeA

Type- TypeB
C

Type- Type- TypeD
E
F

Total

31

0

15

5

0

0

51

42

0

27

19

2

0

90

26

0

3

10

2

0

41

7

4

8

0

2

0

21

8

1

1

2

0

0

12

114

5

54

36

6

0

215

Photo.5 Example of Fatigue Crack at Ohira ViaductMeasurement Plan and Measurement Contents(1) Outline of measurements
SFRC pavement can be applied to reduce deformation of the deck plates of the steel
deck and improve load distribution, thereby diminishing strain localized around weld
areas, in other words local strains. In this study, strain gauges were attached, mainly to190the welds of deck plates, but also to transverse ribs and U-ribs, in order to ascertain theeffect of SFRC pavement in terms for the reduction of local stress on the steel deck.
As the strain reduction effect of SFRC pavement was predicted to differ between strain
gauge attachment location 1 at approximately the middle point between piers P4 andP5, where the bridge's vertical deflection is comparatively large, and strain gauge
attachment location 2 near intermediate support P5, where vertical deflection iscomparatively small, it was decided to perform measurements at these two locations.
The measurement target was the box
Wheel position of survey line agirder on the overtaking lane side
Survey(median strip side), and survey lines alines
to f were selected, taking into
consideration the positions of the
U-ribswheel tracks and U-ribs. It was
assumed that the right wheels would
travel over survey lines a to e, and theleft wheels over survey line f. The
Main girder web
Main girder web
positions of the survey lines are shown
Fig.7 Position of Truck Loading Survey Linesin Fig.7.

191

2160

220 220100
1300

220 220100

220 220100
1300

220 220
100

2070

45

220

1850

220

45

(2) Loading test caseThree-axle trucks with two single wheels at the front and four double wheels at therear end were used for the loading test. The gross weight of each truck was adjusted tobe approximately 200 kN. Fig.8 shows the details of the front wheels, front-rear wheels,and rear-rear wheels. Regarding the loading of the double tires on the rear axes, loading
was done so that the centerline of the double tires coincided with the survey line. As
shown in Fig.8, the loading position of the front wheels in the width direction is 45 mm
retracted from the position of the external wheels. Therefore, the front wheels passdirectly over the middle of the U-ribs, as the positions of the double tires of the rear
wheels span exactly the survey lines. To clarify the loads of the trucks that were used,
the axle loads of the trucks were measured at the time of the asphalt pavement and after
SFRC application. The measurement results are listed below.
3220
1330
1) Axle load of loaded truck at time of
240
240
240
asphalt
Total weight: 222.0 kN
Front wheel axle weight: 58.2 kN
Front-rear wheel axle weight: 83.2 kN
Rear-rear wheel axle weight: 72.8 kN2) Axle load of loaded truck during
SFRC work
Total weight: 222.3 kN
Front wheels
Front-rear wheels
Rear-rear wheels
Front wheel axle weight: 56.4 kN
Fig.8 Details of Truck Wheel Placement andFront-rear wheel axle weight: 87.0 kN
Contact Area
Rear-rear wheel axle weight: 73.3 kN
The differences between the two types of truck were judged to be small, and themeasurement data was sorted under the assumption that trucks of the same weight were
loaded, without taking into consideration axle weight differences.

(3) Measurement schedule
Measurements were performed twice, during asphalt pavement prior to removal(before reinforcement), and about one week following SFRC placement (after
reinforcement), in order to determine the efficiency of SFRC pavement for strain
reduction in the steel deck.
The measurement dates and the average temperature data of the Utsunomiya Local
Meteorological Observatory and Oyama are listed below.
1) During asphalt pavement: 2 October 2007, 19.1°C, cloudy2) During SFRC pavement: 30 October 2007, 16.1°C, cloudy3) During SFRC pavement: 11 November 2007, 14.3°C, cloudy(4) Positions of measurement strain gauges
Strain gauges were used mainly to measure the strain in the deck plates in the box
girders. As shown in Fig.9, the strains along the bridge axis of the deck plates and theaxis orthogonal to the bridge axis were measured in the vicinity of the transverse ribs
(D-D section) and transverse rib span center (E-E section). At the intersection of thetransverse rib and U-rib, a 3-axis rosette gauge was used, taking into consideration thefact that the principal stress direction is unknown, as shown in the lower right part ofthe same figure. Regarding the organization of the strain data, the values of the strain
gauges on axis 1 and axis 2 were used as is, and for axis 3, the principal strain values
were used. Table 3 shows the measuring equipment that was used. For the strain gauges
at the scallop locations, a gauge length of 3 mm was attached at a position 5 mm fromthe weld toe, as shown in Fig.10.
Table3 Used of Measuring Equipment
Equipment Name
1-axis strain gauge：
FLA-3-11-10LTSAStraingauge

2-axis strain gauge：
FCA-3-11-10LTSA

3-axis strain gauge：
FRA-3-11-10LTSA

Data logger
Switch box

192

TDS-303
ASW-50C

Usage
Location
Steel beam
Steel beamTransverse rib

Fig.9 Gauge Positions and Gauge Numbers

Fig.10 Scallop Position Gauges

193

WorksAt the attached steel plate on the steel-deck plate, removal of the existing asphalt
pavement by hands were executed paying attention to the position of bolts based on
preliminary-survey results. In this case, the breaker that had an even edge so thatdamage may not be given to the steel deck plates was used. Much attention was paid
especially so that damage might not be done to the bolt that was binding the attachedsteel plate tight. After removal of the existing asphalt pavement, the water jet machine
stripped off all the water-resistant layers that remained on the deck plate.
As grinding processing in the deck plate to secure an excellent adhesion, the rust andforeign substances was removed by using two or more sets of shot blasting apparatuses.
Small-sized apparatuses (0.35m in width) were used for the corner of steel deck plates,
while large-sized apparatuses (1.0m in width) were used for the central part. In both
sized apparatuses the density of shot blasting using steel balls of 1.4 mm in diameter
was controlled by 300kg /m2. Application of epoxy resin adhesive started by 30
minutes before placing of SFRC and the speed of application was accommodated to theplacing of SFRC.
Concrete with high-early-strength Portland cement and expansive admixture was
manufactured in ready-mixed concrete plant and transported with agitator vehicles tothe site. The steel fiber was put in the agitator car with the special device set up on thesite, and mixed with concrete. The expansive admixture was used in order to reduce
cracking due to drying shrinkage. After slump test and test for air content of the
manufactured fresh SFRC samples was conducted to check the quality, the SFRC wasdischarged.
The SFRC was placed and compacted with a finisher. The SFRC was cured withplastic sheets for one day after placing and sprayed water for the following four days.
Measurement and observations
To clarify the strain reducing effect of SFRC pavement, the results of comparisons
before and after SFRC reinforcement at the strain gauges near deck plates, andcomparisons before and after SFRC reinforcement, in other words at the time of theasphalt pavement and at the time of the SFRC pavement (hereafter, before and after
SFRC reinforcement), at the strain gauges at the cross beams were compiled and sorted.
The results are shown in Figure.11. The strain amplitude, which is the difference
between the maximum strain and minimum strain during the passage of the trucks usedin assessing fatigue durability performance, was used as the strain. The two-digit
numbers and three-digit numbers in the strain gauge column indicate mid-span and P5
pier vicinity strain gauge attachment positions, respectively. Regarding thecomparisons before and after SFRC reinforcement, the ratio (%) of the value after
SFRC reinforcement (100 × value after reinforcement ÷ value before reinforcement)
was used, defining the value before SFRC reinforcement as 100, and this numericvalue is indicated at the right end of the strain amplitude (bar graph) in Fig.11.
Particularly in the case of deck plates (shaded parts in Fig.11), this value was often
10% or lower.
Looking at the strain values of the shaded parts, none of the gauges whose strain
amplitude caused by wheel loads was on the order of 400 to 500 µ (84 to 105N/mm2)
at the asphalt stage reach 100 µ (21 N/mm2).
Moreover, regarding the welds of the deck plates and transverse ribs (E and F in

194

Fig.11), there are many gauges withGa
values after SFRC pavement that have Symbol ug 0edropped to as low as 20% to 30% of4749
the original value, a result showing
50
A
1
that SFRC pavement is highly
147
149
effective in reducing fatigue cracks in150
53
the steel deck.
Conclusions
SFRC pavement was laid over alength of 213 m consisting of one
3-span continuous 2-box girder andtwo simple plate box girders of theinbound (Takasaki-bound) lanes of
the Ohira Viaduct. Before and after
reinforcement of the 3-spancontinuous 2-box girder that was
selected for strain measurement, inother words when the pavement
consisted of asphalt and SFRC,
respectively, loading tests were
performed. Trucks were used as theloads, and the reduction in strain
obtained through SFRC pavement
prior to opening of the bridge to traffic
was investigated.
The main findings are summarized
below.
1) The strain reducing effect at deckplates locations is extremely large,
particularly at the welds between deckplates and U-ribs, and in some cases,
the strain value was reduced to as little
as 1/10th the level following SFRC
reinforcement.
2) Even at the welds between
transverse ribs and U-ribs, the strain
value was reduced to between 1/2 and1/3 as the result of SFRC
reinforcement, clearly attesting to thestrain reducing effect of SFRC
reinforcement, though to a lesser level
than at deck plates.

A2

B
C
D
E1

E2

Ｆ1

Ｆ2

G1

G2

Symbol
A1

54
55
153
154
155
51
151
56
156
60
160
2
5
102
105
14
17
20
23
114
117
120
123
7
9
10
107
109
110
25
26
27
125
126
127
38
39
40
138
139
140
30
33
36
130
133
136

Strain Amplitude (μ)
100

200

300

400

500

600

9
8
8
6
6
12
11
18
28
12
18
18
12
15
27
30
34
39
15
14
12
10
67
59
66
86
53
36
66
77
7
8
11
8
7
8
32
65
41
61
13
31
32
29
29
26
28
27
30
37
35
29
28
32

Before reinforcement
After reinforcement

Point of Focus

Strain Direction

A2

Weld between deck plate and U-rib (U- Strain of deck plate
rib span)
Vertical strain of U-rib

B

Weld between deck and vertical stiffene Strain of deck plate

C

Weld between main girder and vertical s Vertical strain of main girder

D

Stud weld of U-rib

E1

Weld between transverse rib and U-rib Strain of deck plateor deck plate3-axis strain of transverse rib

E2
F1
F2
G1
G2

Weld between deck plate and U-rib(transverse rib crossing)
Weld between transverse rib and U-rib

Strain of U-rib

Strain of deck plate
Vertical strain of U-rib
3-axis strain of transverse rib
Vertical strain of U-rib

Fig.11 Strain before and after SFRC Reinforcement

195

FINAL REMARKSThe measurements conducted in the present study were carried out immediately after
reinforcement, when the reinforcing effect of SFRC can be expected to be greatest, anda large strain reducing effect believed to contribute to the improvement of fatigue
durability of steel decks was observed. Even after the lapse of approximately fourmonths under service following SFRC placement, no cracks were observed at thesurface of the SFRC, and the placement of CFRP grids as a countermeasure [10] toprevent cracks in main girder webs is also believed to have an effect, but theoccurrence of cracks to some extent in the future is still considered possible.
Follow-ups will be done and the effect of the SFRC pavement that was placed this time
as a fatigue countermeasure for the steel deck, the persistence of this effect, thedurability of the pavement proper, and other aspects will be examined.

REFERENCES
1. Nishikawa, Steel deck surfacing with SFRC—A New Relationship between Steeland Concrete, Bridge and Foundation Engineering, August 2005, pp. 84-87.
2. Yamada, Fatigue Damage of Steel Plate Decks under Heavy Traffic—Cases around
Nagoya, Japan Society of Civil Engineers, 10th Symposium on Steel Structures andBridges. August 2007.
3. Satoh et al., Composite Effect of Steel Deck Surfaced with Steel Fiber Reinforced
Concrete Pavement, Steel Structures and Bridges, February 1986, pp. 26-30.
4. Kagata et al., Measures against Fatigue Damage of Steel Decks through SFRC
Paving—The Case of Paving Work on National Highway 357's Yokohama Bay
Bridge, Steel Structures and Bridges, October 2004, pp. 27-32.
5. Yamada, Fatigue Reducing Effect of SFR Pavement on Steel Decks, 62nd Annual
Conference of the Japan Society of Civil Engineers, September 2007.
6. Kikuchi et al., Fatigue Countermeasures for Steel Plate Deck of Shonan-Ohashi
Bridge, Japan Society of Civil Engineers, 10th Symposium on Steel Structures andBridges. August 2007.
7. Kodama et al., Fatigue Countermeasures for Steel Decks using SFRC during
In-Service Period, Bridge and Foundation Engineering, November 2006, pp. 30-38.
8. Kogoshi et al., Measurement of Effect of SFRC Reinforcement of Existing Steel
Decks of Bridges," 62nd Annual Conference of the Japan Society of Civil Engineers,
September 2007.
9. Ono et al., "Water Immersion/Wheel Load Fatigue Test for Steel Deck Surfaced with
Steel Fiber Reinforced Concrete, 62nd Annual Conference of the Japan Society of
Civil Engineers, September 2007.
10. Hitose et al., A Study on Rebars for Increasing Durability of SFRC Pavement on
Steel Plate Decks, 62nd Annual Conference of the Japan Society of Civil Engineers,
September 2007.

196

RAPID BRIDGE REPAIR / REHABILITATION IN WASHINGTON STATE
Jugesh Kapur 1
Abstract
The Washington State Department of Transportation (WSDOT) used a very costeffective technique using Carbon Fiber Reinforced Polymer (CFRP) to rehabilitateand strengthen a deteriorated precast concrete bridge. The beams on this bridge had
experienced spalling of concrete and corrosion of reinforcement. WSDOT also
replaced four, 122 cm movement steel modular joints on an Interstate 90 floating
bridge due to fatigue cracking of the transverse center beams in the modular joint. In
order to meet the new fatigue resistant design criteria, the new center beams are solidand 2.5cm wider than the center beams in the original modular expansion joint. Bothprojects were done very efficiently and rapidly without loss of quality.
US 2 Ebey Slough Viaduct – Repair with Carbon Fiber Reinforced Polymer(CFRP)The Washington State Department of Transportation (WSDOT) has
effectively used Carbon Fiber Reinforced Polymer (CFRP) to rehabilitate portions ofa deteriorated precast concrete bridge located near Everett, Washington at beginningof highway US 2. Portions of the bridge were repaired in 1999 and 2007 with the final
section due to be completed in 2010. CFRP strengthening of vertical load carrying
elements is relatively new in the United States even though it is used extensively inEurope and Japan. CFRP strengthening can be a cost-effective alternative toreplacement or posting of structurally deficient bridges. Despite its high cost, CFRP
possesses high strength and is light in weight. It is also corrosion resistant, easily
constructible and has a high fatigue resistance. These qualities make it a favorable
material for the rehabilitation of our nation’s transportation infrastructure.
Existing Bridge
The Ebey Slough Viaduct was constructed during the late 1960’s and opened
to traffic in 1968. The bridge carries nearly 35,000 vehicles per day and provides avital link between western and eastern Washington. The bridge spans a river andslough area that floods every year. The bridge has several different superstructure
types and is 11,909 feet (3630 meters) in length with a total of 256 spans.

EBEY SLOUGH VIADUCT

1State Bridge Engineer, Washington State Dept. of Transportation, Olympia, WA

197

A significant portion of the bridge, 226 spans – 7,410 feet (2259 meters) has
conventionally reinforced precast concrete double tee concrete units, also known as
inverted tub units. The concrete unit spans approximately 38 feet (11.6 meters)
between piers with each span composed of six superstructure units bolted together
transversely through their webs.
Each individual unit has a bundle of four
#11 reinforcing bars at the bottom of each web
covered with 1 inch of concrete. The precast
concrete units also have grouted keyways between
them. Over the years the grout in the keyways had
deteriorated allowing water and deicing salt topenetrate between the precast units which caused
the main longitudinal steel reinforcing bars tocorrode and spall off the concrete cover. Areas of
exposed and corroded reinforcing steel are scattered,
but mostly predominant in the exterior units.
WSDOT contracted with Construction Technology Laboratories to analyze
several randomly obtained core samples with the objective of evaluating the technicalfeasibility of repairing this structure. An analysis of the concrete core samples
determined that structural rehabilitation was feasible.

198Repair and Rehabilitation Using CFRP
WSDOT bridge engineers reviewed different options and decided to repair the
precast tub units using CFRP. Since the amount of section loss in the steel reinforcing
was unknown, the CFRP would make up for the loss. CFRP material properties andapplications methods differ from conventional bridge materials and how these

realities affect typical civil structural engineering design philosophies. The carbon
fiber has an elastic modulus comparable to that of steel and an ultimate strength that ismuch greater than that of steel. Carbon does not possess the same ductility. This isdifferent with respect to the traditional ductile concrete and reinforcing steel design
philosophy.
CFRP composites have positive
attributes including high strength-to-weightratios. Additionally, the designer is afforded theflexibility of orienting stiffness and strength in
the directions required. FRP composites havehigh resistance to fatigue and corrosion.
Compared to other rehabilitation methods, FRP
composites minimize handling difficulties andcan be installed in limited access situations,
often with minimal disruption to traffic.
In spite of these advantages, many perceived technical and institutional
barriers still exist to FRP composite usage in the civil engineering arena such as,
standardized/codified design criteria is limited and there is a lack of construction
specifications available. Durability is another issue. The aerospace industry has
conducted extensive durability tests on FRP composites relevant to their needs.
Obviously, the applicability of some of these tests to civil engineering applications is199somewhat contentious. Plausible durability tests include resistance to humidity, salt
water immersion, alkali, dry heat, fuel, ultraviolet weathering, and freeze/thaw
cycling. The questions which need to be definitively answered are: Which of thesetests are pertinent and should be required? For what durations should such tests be
conducted?
FRP composite costs are perceived to be high. In rationalizing material costs,
other factors must be considered: life cycle costs, traffic disruptions, durability,
worker safety, and real estate costs associated with wholesale bridge replacement.
One important issue was the sequencing under which carbon fiber plies are bonded toload carrying elements. Manufacturers and suppliers of CFRP materials have asserted
that the plies can be bonded to structural bending elements of a bridge while thatstructure is under full traffic. Structural engineers disagree on the effectiveness ofstrengthening performed under full live load.
The main concern is that imposed live load will subject both the resin bonding
agent and the carbon fiber plies to micro-strain during the curing process. Assuming
full development of bond to the substrate structural material, the carbon fiber plies
will augment the ultimate capacity of the structural bending element. The carbon
fiber plies will not participate in resisting future live loads that are less than those thatwere imposed during the resin curing process. This could have potential long-term
durability implications depending upon the condition of the structure before
rehabilitation, the live load levels anticipated during bond cure, and the futureenvironment to which the structure will be subjected.
Prior to structural design of the Ebey
Viaduct repairs, a field survey was conducted toestimate steel corrosion losses at various
locations. Based upon these estimates, a stressstrain compatibility analyses was performed toestablish the carbon fiber ply cross sectional area
required to regain structural moment capacity.
This was done for varying levels of steel
reinforcement bar corrosion loss.
Carbon fiber ply application plans specifying carbon fiber ply tensile capacity
requirements were developed for each span for the contract documents. Multiple
longitudinal plies where required by the design calculations. In areas where very
severe spalling had occurred, transversely oriented carbon fiber plies were specified toresist horizontal shear across the existing concrete-to-repair mortar interface.
Epoxy injection of all cracks greater than 0.02" (0.51 mm) wide was required.
Concrete surface preparation follows. In addition to removal of spalling concrete, thisincludes application of passive grout to all exposed steel reinforcement following its
sandblasting to a white metal finish.

200

All existing sharp edges/corners in the concrete are ground smooth. Repair
mortar placement follows. Once the repair mortar cures, a primer is applied to thebottoms of all concrete webs to prepare the surfaces for carbon fiber ply bonding.
Putty is then used to fill all surface concavities. A resin undercoat is then applied towhich the carbon fiber plies are immediately bonded. After the plies are bonded androlled smooth, a resin overcoat is applied. Sand was broadcast onto the resin overcoat
for aesthetic reasons and to provide some level of resistance to ultraviolet
degradation.
Performing either repair mortar placement or carbon fiber bonding work underfull traffic conditions was ruled out due to the high amount of truck traffic on the
structure. A compromise was reached by electing to perform these repairs during
either full traffic closure or under restricted traffic conditions. Restricted traffic wasdefined as only vehicles under 5 tons (4546 kg) at speeds below 25 miles (40 km) per
hour. This gave four different combinations that determined how to perform the workon the remainder of the bridge.

Pilot Project on eight spans
WSDOT bridge engineers selected
a 304 feet (92.7 meters) long eight span
section of the bridge for a pilot project.
The objective of this pilot project was toassess the feasibility of rehabilitating all
the precast concrete tub units on the entire
bridge using advanced composite CFRP
materials, both from an economic and a
durability perspective. The use of CFRP
would provide a way to repair andstrengthen the bridge with minimal
disruptions to traffic.
The work began in June 1999 and was completed by August 1999. The
contract was awarded to Concrete Barrier Incorporated for nearly $970,000 with 40
days to complete the work. Mitsubishi Chemical supplied the carbon fiber system.
The overall scope of work included setting up a containment and work
platform. Removal of the existing ACP overlay to allow repair of deck delaminations
and to inspect and repair the longitudinal keyway joints between adjacent concrete
units. To repair the grouted keyways, the contractor sawcut 9 inches (23 cm) deep tothe mid-depth of the keyways and cleaned and filled the resulting 1 inch wide gap
with low modulus epoxy. A new deck membrane and ACP overlay was then placedon the roadway deck side of these eight spans. Upon completion of the overlay
repairs, the spalled concrete web surfaces were prepared for placement of repair
mortar. After repair mortar placement, carbon fiber plies were bonded to the web
bottoms.

201

Corrosion inhibitor was applied to all exposed surfaces of the precast concrete
units to enhance durability. The overall rehabilitation strategy involved the removal ofdeteriorated concrete adjacent to spalls, blast cleaning of all exposed corroded steel toa white metal condition, epoxy injection of cracks wider than 0.02", repair mortar
placement, miscellaneous concrete surface preparation, and bonding of CFRP strips to
the web bottoms. Structural design calculations were set up in an automatedelectronic format to establish the cross sectional CFRP requirements for varying
levels of longitudinal steel reinforcement corrosion
loss.
The carbon fiber supplier's willingness to beon the jobsite to provide technical guidance and train
to the contractor was directly related to the success ofthe project. The supplier and contractor's work crew
had a meeting prior to the carbon fiber application.
Previous projects nationally have demonstrated that
CFRP can be bonded under full traffic conditions tostrengthen bridge superstructures. In order to evaluate
several construction staging options, span-specifictraffic control requirements with respect to placementof repair mortar and CFRP strip bonding were
specified. Two traffic control conditions were
imposed during construction: full traffic closure and a restricted traffic condition withvehicle weights limited to five tons at a maximum speed of 25 miles (40 km) per
hour. Durability and performance of spans repaired under the two different traffic
control scenarios will be examined and taken into consideration in developing a longterm repair strategy for the entire bridge.
Total cost of this initial pilot project was
about $970,000 which equates to about $106/SF($1141/m2) of bridge deck area. This cost is
10% or less of the cost of bridge replacement. A
breakdown of seven bid items representing
approximately 80% of the contract cost shows
the removal and replacement of concrete
accounted for about 40% of the total with the
CFRP application accounted for about 8% of thetotal cost.
Ten years following construction, all
CFRP repaired concrete areas have been inspected with the longitudinal joint repairs
and carbon fiber reinforcement performing as originally designed with no visible
deterioration. It is anticipated that this repair will provide 25 years or more of service.
Phase 2 Project - Rehabilitate 128 spans

202

WSDOT bridge engineers identified 136 spans of the tub units for the second
phase of the Ebey Slough Viaduct rehabilitation. The project followed the same
design as specified in the 1999 pilot project. The contract was awarded to Wilder
Construction Company for nearly $7.9 million with 97 days to complete the work.
The work began in April 2007 and was completed by November 2007.
The contract plans specified that all traffic be restricted during the applicationof the carbon fiber which required the use of the parallel bridge at times to carry both
directions of traffic. The contractor was allowed to determine the number of layers ofcarbon fiber to apply based on the amount of corrosion of the steel reinforcing bars.
This provision became a problem during the project since WSDOT review ofassumptions made by the contractor did not justify the number of layers of carbon
fiber the contractor applied. The total cost of the carbon fiber material was nearly$1.5 million or 19% of the total project cost.
Two years following construction, all CFRP repaired concrete areas have beeninspected with the carbon fiber reinforcement performing as originally designed withno visible deterioration. It is anticipated that this repair will provide 25 years or moreof service.

Phase 3 Final Project - Rehabilitate 82 spans
The final 75 spans of the precast concrete tub units for the Ebey Slough
Viaduct are scheduled to be rehabilitated in 2010. The project will follow the same
design as specified in the 1999 and 2007 projects. This project has a budget of nearly$31 million. It is anticipated that the actual cost will be substantially less that what iscurrently budgeted.
The traffic control and construction access will be more difficult than the previous
two phases. These bridge spans are over water so the contractor must install work
platforms from the existing bridge deck requiring additional lane closures.
Connections to the parallel bridge are also more difficult in this area thus the traffic
control options will be more complicated.

I-90 Homer Hadley Floating Bridge – Replacement of Modular Expansion Joints
The Interstate 90 “Homer M. Hadley” bridge is thethird floating bridge to be constructed across Lake
Washington between Mercer Island and Seattle and carries
five lanes of traffic (three for westbound and two reversible).
This bridge was officially opened on June 4, 1989 and is
5,811 feet (1771 meters) in length with eighteen floating203pontoons. The three westbound lanes have an Average Daily Traffic (ADT) of
60,000 and the two reversible lanes have an ADT of 12,000.
The bridge has transition spans at both ends that allow traffic to move from theapproaches to the floating sections and to allow boats to go under the bridge. The
longitudinal movement, which represents a combination of wind and wave loading,seasonal water level fluctuations, and thermal variations in the bridge areaccomplished through the steel modular expansion joints.
ORIGINAL MODULAR EXPANSION JOINTSThe section of the bridge with the threewestbound lanes has two large movement
modular joints (48 inch, 122 cm) that are 63
feet ((19.2 meters) in length and the section
with two reversible lanes has two large
movement modular joints (48 inch, 122 cm)
that are 40 feet (12.2 meters) in length. Thus
there are a total of four of these sizes of joints
on the bridge.
The original modular joints were designed, fabricated, and installed prior to
the advent of fatigue-based design and manufacturing criteria for modular expansion
joints. The joint manufacturer chose to use a hollow tube shape for the maintransverse center beams.
75mm

125 mmFatigue cracking initiated in
the hollow tube and the welded
support bar connection within 18
months after the Homer Hadley Bridge
opened to traffic.

WSDOT developed someresearch projects with the Universityof Washington to investigate the causeof the cracking. The university
performed field measurements of the
dynamic wheel loads. The location and extent of the cracks were also carefully
monitored. Repairs were made in several instances by welding the cracks but nearly
all of these locations cracked again.
Fatigue Design of New Modular Expansion Joints
In November 1990, the pontoons in the parallel original I-90 Lacey V.
Murrow floating bridge, which was being rehabilitated at the time, sunk. WSDOT
recognized the need to improve on the design of the large movement modular joints
during the design of the I-90 Lacey V Murrow replacement.

204

As part of that effort, WSDOT worked with D.S. Brown, the joints'
manufacturer, the German-based Maurer Sohne Company, and University of
Innsbruck (Austria) to develop the first generation fatigue resistant design
specification for WSDOT. This new design was used on the new Lacey V. Murrow
Bridge replacement.
Following the work by WSDOT, there was a substantial amount of interest
throughout the United States to improve the fatigue resistance of steel modular
expansion joints. The Transportation Research Board (TRB) subsequently funded
NCHRP Project 12-40. This project culminated in the publication and release of
NCHRP Report 402, which provided comprehensive fatigue-based test and design
specifications for modular expansion joints.
To date, all of the modular expansion joints designed and installed based on
this new fatigue design criteria are performing well.
Project Planning and Design
A decision was made in 2007 that all four of the 48 inch (122 cm) movementsteel modular joints needed to be replaced due to the amount of fatigue cracking in thetransverse center beams and the high risk associated with a potential failure. In order
to meet the fatigue resistant design criteria, the new center beams are solid and one
inch wider than the center beams in the original modular expansion joint.

180 mm

100mm

The replacement of the joints required WSDOT to determine the best way forthe joints to be manufactured, delivered and installed. WSDOT developed an initial
estimate and plan then had a meeting with the Association of General Contractors(AGC) to review the details.
At the meeting, the AGC and WSDOT discussed some of these details:

205

1. The amount of time to replace the joints – three weeks for the westboundlanes and 3 weeks for the reversible lanes.
2. The joints should be replaced in two separate closures, one for the
westbound lanes and one for the reversible lanes.
3. WSDOT should purchase the joints and advertise thereplacement/installation project separately.
4. Make sure the liability issues and warranty issues are understood andaddressed.
WSDOT decided to purchase the joints directly from the manufacture andhave them shipped to Washington State. These joints would be handed over to thecontractor that was given the joint replacement contract. WSDOT supplying the joints
would speed up the manufacture and delivery time, eliminate contractor mark-up,
reduce contractor risk for acceptance of the joint. Having the joint design prior to thebid will allow WSDOT to fully design the replacement reinforcement and dowels into
the block out area. The contractor would then have a better idea of the actual
removal-installation details.
Project Construction
The expansion joint contract
was awarded to the General
Construction Company for $5.7
million with the final construction cost
being $5.85 million. The 4 modular
expansion joints were replaced in twostages.
Stage 1 replaced the two 40
foot long joints located in thereversible lane roadway. All traffic
used only the three westbound lanes. The first closure began on May 4, 2009 and wascompleted on May 16, 2009. The contract allowed 19 days with the contractor
completing the work in 12 days. For the first closure, the contract had a provision to
pay the contractor an extra $40,000 for each 24 hour period completed early. Total
was not to exceed $160,000.
The first closure and replacement of the two 40 foot (12.2 meters) long joints
allowed the contractor to better understand the work. We thought that the lessons
learned during the May work would enhance the contractor's productivity during themore complicated stage 2 closure.
Stage 2 replaced the 63 foot (19.2 meters) long joints located in the westboundlane roadway. All traffic had to use the two reversible lanes. The second closure
began on July 5, 2009 and was completed on July 17, 2009. The contract allowed 21days with the contractor completing the work in 12 days. For the first closure, the206contract had a provision to pay the contractor an extra $80,000 for each 24 hour
period completed early. Total not to exceed $320,000.
.
Our Special Provisions required thatthe contractor submit structural analyses
verifying the load carrying capacity of thebridge for the specific cranes that would be
used. Given the weight of each joint being
installed and the boom radii that would be
required to position these joints in theirrespective blockouts, we realized that thecontractor would need some pretty substantial
lifting equipment. Additionally, we requiredthat the contractor operate all cranes on a
separate pad and load distribution system designed in accordance with AASHTOLRFD Bridge Design Specifications.
The contractor used a 250-ton (228,000 kg) and a 300-ton (272,000 kg) crane
operating on the respective transition spans to install joints on the reversible lane
roadway. The contractor used a barge-mounted crane to install the two modular joints
on the westbound lane roadway.
The contractor developed a good
method to remove the concrete in theheader area on both sides of the existing
expansion joint and removal of theexisting joint. A large concrete saw wasused to cut a line across the bridge at theremoval limits. Workers then usedchipping hammers to remove theconcrete.
The first of the technical issues
was the fact that the overall width of thereplacement modular joints was greater than that of the originally installed joints.
This created some issues with the traffic barriers. When a new bridge is constructed,
blockouts are formed in the concrete deck or edge beam to accommodate the modular
expansion joint later on. After the modular joint is installed, the concrete barriers areconstructed up to the outside limits of the modular joint with steel embedment for
attaching steel sliding cover plates.
Removing and reconstructing the ends of each concrete barrier would haveadded at least three days to the total construction duration. In lieu of demolishing theends of the existing concrete barrier it was decided to install all four joints incontracted configurations before expanding them prior to anchorage and concrete
placement. The bottoms of the concrete barriers were chipped out to accommodate
the edge beam upturns beyond the curb line.

207

The next technical issue was the gap to
set the replacement modular expansion joint inplace. The cumulative gap setting is primarily afunction of wind and wave motion and waterlevel, not the temperature of the bridge, thefloating portion of which is moderated by thetemperature of the lake water. It was decided to
set reference pins in the roadway slab prior to
removal of the existing joints, then to install thereplacement joints so that they would have the same cumulative gap setting as theoriginal joints.
The overall project was considered a success. It is anticipated that the lessons
learned on this project will be used when WSDOT replaces other large movement
bridge expansion joints in the future.

208

25th US-Japan
Bridge Engineering Workshop

Session5

Maintenance
Development of an Expansion Device for Cold Regions
By Shinya Omote, Hiroshi Mitamura, and Hiroaki Nishi
Fatigue and Corrosion in Concrete Decks with Asphalt Surfacing
By Yoshiki Tanaka, Jun Murakoshi, and Yuko Nagaya
Replacing The Suspender Ropes of a Tied Arch Bridge Using Suspension Bridge Methods
By Barney T. Martin and Blaise A. Blabac
Suspension Bridge Cables: 200 Years of Empiricism, Analysis and Management
By Bojidar Yanev

Page Left Blank

DEVELOPMENT OF AN EXPANSION DEVICE FOR COLD REGIONS- TEST INSTALLATION OF AN EXPANSION DEVICE ON A
NATIONAL ROUTE 274 INTERCHANGE VIADUCT Shinya Omote1
Dr. Ing.Hiroshi Mitamura2
Dr. Ing.Hiroaki Nishi3

Abstract
Bridge expansion devices in Hokkaido face severe conditions (including frost/saltdamage and impact from snow-removal activities), meaning that they deteriorate andsustain damage faster than other bridge members． As the replacement of damaged
expansion devices has increased dramatically in recent years, a model with high fatigue
durability and load capacity was developed to reduce future life cycle costs. The
expansion device designed for cold regions was installed on a bridge administered by
the Sapporo city to examine its workability, load-carrying capacity, waterproof
performance and anticorrosive function, as well as to enable on-site verification
through data accumulation, including information on the impact of snow-removal
activities and stress propagation characteristics under conditions with runningvehicles.

Introduction
Bridge expansion devices in cold, snowy regions operate in severe conditions(including frost/salt damage and impact from snow-removal activities), meaning thatthey deteriorate and sustain damage faster than other bridge members. The replacement
frequency of damaged expansion devices has increased dramatically in recent years.
This paper presents the results of a study on the development and installation of an
expansion device with high fatigue durability and load capacity aimed at reducing
future life cycle costs.

1 Results of a survey on the conditions of expansion devices

1

Researcher, Structures Research Team, Public Works Research Institute, Civil
Engineering Research Institute for Cold Region
2
Senior Researcher, Structures Research Team, Public Works Research Institute, Civil
Engineering Research Institute for Cold Region
3
Team Reader, Structures Research Team, Public Works Research Institute, Civil
Engineering Research Institute for Cold Region2111.1 Survey overview
The targets of the survey were 561 sections with expansion devices and 186 ridges
on heavily trafficked major arterial roads around Sapporo, as shown in Fig. 1.1. ofthese, 136 bridges and 459 sections with expansion devices were on national highways.
In the field survey, the damage conditions of bridges and expansion devices were
examined mainly through visual observation.

FIG. 1.1 SURVEY LOCATIONS
1.2 Status of bridges and existing expansion devices(1) Status of bridges
River bridges accounted for approximately 80% of the total, while the ratios of
viaducts and bridges over railways were a little less than 10% each.
(2) Status of expansion devices
Figure 1.3 shows the types of target expansion devices, of which 60% were theload-support types, 20% were the no-joint type and the other 20% were the match
type.

FIG. 1.3 TYPES OF EXPANSION DEVICE
1.3 Age occurrence conditions
Figure 1.6 shows the results of the survey on damage to expansion devices.
(1) Load-support type(a)Unevenness of face plates (Photo 1), insufficient expansion gaps (Photo 2) andcuts made by snowplows (Photo 3) were observed in many cases with the212steel-finger type, and sediment accumulation, clogging and corrosion were foundat water cutoff parts.In cases using the beam type, differences in beam level anddeformation were observed (Photo 4). At water cutoff parts, blockage of
expansion gaps with sediment (Photo 5) and loss of cutoff rubber (Photo 6) were
found.
(b)Extrusion of sealing material (Photo 7) was observed in many cases withhorizontal simple-installation-type expansion devices made of steel, although
there was little damage to their main bodies. In cases with vertical-type devices,
deformation (Photo 8), breakage and corrosion of the main bodies were significant,loss of waterproof ness was observed and water leakage was severe.
(c)In all cases with the load-support type, scaling of the pavement around expansion
devices (Photo 9) and hollowness, scaling and wear of site-placed concrete (Photo10) were observed. Such damage was significant on the front side in the side
facing the direction of traffic.
(2) Match type(a)Breakage of steel beams (Photo 11) and other serious damage to main units were
found in some cases with the beam type. Blockage of water cutoff parts withsediment and loss (Photo 12) and cracking (Photo 13) of cutoff rubber were also
observed. In cases with devices made of rubber, peeling of surface rubber was
seen.
(b)Scaling of the pavement around expansion devices (Photo 14) and hollowness,scaling and wear (Photo 15) of site-placed concrete were observed in all types.
(3) No-joint type(a)While the status of internal structures was unknown because the survey was
conducted through visual observation only, cracking (Photo 16), loosening,scaling, unevenness, caving and other damage to pavement were observed in thevast majority of cases.

FIG. 1.6 CONDITIONS OF DAMAGE TO EXPANSION DEVICES

213

(4) Other(a)Many cases of curb breakage and edge-plate breakage/deformation (Photo 17)
were observed at wheel guard parts, while deterioration and peeling of sealingmaterials and blockage with sediment were seen at the water cutoff parts ofexpansion gaps.
(b)Significant breakage of main units (Photo 18), rusting, corrosion and other serious
damage were observed in some parts of sidewalk areas. Scaling and loosening ofsurrounding pavement and damage to site-placed concrete were also found in all
types.
2 Estimation of the causes of damage
Possible causes of damage to expansion devices peculiar to Hokkaido are listed
below.
(1) Damage caused by snowplows
Scaling of pavement and concrete around expansion devices was thought to result
from contact with snowplows during snow removal work and from fatigue causedby the wheel loads of large vehicles.
(2) Damage caused by sediment and compacted snow
Breakage and loss of cutoff material was thought to result from the transmission ofwheel loads to expansion gaps due to blockage with sediment and compacted snow.
(3) Damage caused by insufficient performance at low temperatures
Extrusion and loss of cutoff rubber was thought to be a result of adhesive failurecaused by the corrosion of steel materials and insufficient performance at low
temperatures.
Scaling and loosening in cases with the no-joint type were thought to be a result of
repeated beam expansion and shrinkage due to temperature variations andinsufficient load-following capability in the expansion devices themselves.
(4) Damage caused by spreading of antifreeze agent
Rusting of steel material and scaling/wear on concrete were thought to result fromaging deterioration and the acceleration of rusting and corrosion from the spreadingof antifreeze agent on road surfaces.
3 Improvements to expansion devices for cold regions
3.1 Performance requirements in cold-region environments
To satisfy performance requirements in the conditions of cold, snowy regionswith wheel loads and aging deterioration, the following considerations are desirable for
expansion devices:(1) Safety against snowplow-related impact must be ensured.
(2) The cutoff structure must be strong enough to resist damage from sedimentaccumulation and compacted snow in expansion gaps where wheel loads aretransmitted.
(3) Resistance to corrosion resulting from the spreading of antifreeze agent, etc. must
be improved. Examination of the internal structure of expansion devices Figure 3.1
presents the examined part, and Photo 3.1 shows the improved expansion device.

214

(4) Installation structure of snowplow inductive plates
To decrease the impact of snowplows, the installation intervals of inductive plates
were reduced from 300 to 225 mm. This structure was adopted to control
snowplow-related damage and disperse wheel loads.
(5) Installation of dust-proof material and slide plates
To prevent the blockage of expansion gaps with sediment or snow and avoid damage
to cutoff materials when wheel loads are transmitted, the entry of sediment was
inhibited and the indentation force of compacted snow was supported by placing
dust-proof material (polyethylene foam) under face plates and laying slide plates
under it. This structure was used to maintain the durability of the elastic sealant usedas the cutoff material.
(6) Improvement of anti-corrosion performance
The anti-corrosion performance of expansion devices was improved by spraying
Al-Mg (aluminum-magnesium) plasma on the steel materials used (the inside of themain units, the face plates, the sides of the main units on the pavement surface andthe inductive plates on the pavement surface). The metal spraying method involves
the application of a layer of aluminum/magnesium (95:5) approximately 100
microns thick melted with plasma onto metal surfaces after Type 1 surface
preparation. The adhesion of sprayed metals can reach more than four times (around
7 N/mm2) that of paint. In a combined cycle test, no corrosion was observed evenafter 6,000 hours (representing more than 100 ordinary years of service), and aself-healing, corrosion-inhibiting effect from the eluted magnesium was observed atcross-cut sections1).

FIG. 3.1 CROSS SECTION OF THE
INTERNAL STRUCTURE OF AN
EXPANSION DEVICEPHOTO 3
IMPROVEDEXPANSION

4 Performance evaluation of the improved expansion device for cold regions
A full-scale specimen was produced for fatigue experiments to evaluate theperformance of the improved expansion device. The specimen had a maximum
expansion gap and a difference in level with the road surface. A wheel runningmachine was used for the experiment.

215

4.1 Running test
(1) Purpose
This experiment was conducted to clarify the process of damage to expansion
devices, and to confirm and summarize the mechanism behind the transmission ofwheel loads to such devices and subsequently to the concrete in the installation
section. The results of the transmission mechanism investigation can also be used toclarify the damage process.
(2) Wheel running machine overview
In the fatigue experiment, the method of using a crank-type wheel running machine(owned by the Civil Engineering Research Institute for Cold Region) was adopted.
This technique has been established as a method of testing fatigue durability
acceleration for road bridge slabs. The experiment is currently under way using amachine with rubber tires rather than the conventional iron wheels (as shown in
Photos 4.1 and 4.2) to simulate actual running conditions.

PHOTO 4.1 WHEEL RUNNING MACHINE(FULL VIEW)

PHOTO 4.2 WHEEL RUNNING
MACHINE

4.2 Running test on an actual bridge(1) Test construction overview
For a running test on an actual bridge, an
expansion device for cold regions was installedat the section shown in Fig. 4.1. The bridge wasan IC viaduct (957 m in length) constructed in
November 1981.The test construction is
expected to clarify the impact of snow removal
in winter and the stress propagation of actualloads applied by many types of large vehicle, as
well as to verify water cutoff and freezing
conditions.

216

Construction site
IC viaduct

FIG. 4.1 LOCATION OF THE
TESTCONSTRUCTION SITE

(2) Overview of the specimen and measurement
As shown in Fig. 4.2, vibration accelerometers (in the vertical and horizontal
directions) and strain gauges (weld gauges) were attached to the specimen. The
actual running tests are currently under way.

Start
Cutting
Removal of existing
expansion device andconcrete
Installation of expansion
devicePlacement of ultra-rapid
hardening concrete

CuringAsphalt pavement

FIG. 4.2 SCHEMATIC DIAGRAM OF
THE SPECIMEN

FIG.4.3 CONSTRUCTIONPROCEDURE(3) Installation conditions
The expansion device was installed following the procedure shown in Fig. 4.3.
Since a main arterial road was being used, the work was conducted at night.
Installation on the main part of the route was conducted separately for the insideand outside lanes, and that on the ramp lanes was conducted simultaneously bycompletely closing the section to traffic. The steps of the installation process canbe summarized as follows:
1) Cutting: Cutting was conducted by marking the predetermined positions (Photo 4.3).
2) Removal of the existing expansion device and concrete: Existing concrete wasremoved using a breaker. The existing expansion device was removed bygas-cutting the finger parts of the face plates (Photo 4.4).
3) Installation of the expansion device: The expansion device was installed in therectangular opening by setting a hanging angle after temporary assembly (Photo
4.5).

217

4) Placement of ultra-rapid hardening concrete: Ultra-rapid hardening concrete wasplaced after installation of the expansion device was complete. The concrete wasblended and mixed on site using a jet mixer truck. After placement, a curing time ofthree hours was observed in principle (Photo 4.6).
5) Asphalt pavement: After the ultra-rapid hardening concrete had cured, its top s
urface was paved with asphalt. While concrete is usually placed up to theinstallation surface in repair work due to time constraints, asphalt pavement wasused in this case in consideration of preventing damage to the surrounding pavement(Photos 4.7 and 4.8).

PHOTO 4.3 CUTTINGPHOTO 4.4 REMOVAL
OF EXISTING DEVICEPHOTO 4.5
INSTALLATION OF NEW
DEVICEPHOTO 4.6 CONCRETE
PLACEMENTPHOTO 4.7 PAVEMENTPHOTO4.8AFTERCOMPLETION

Conclusion
The survey results indicated that the most important points were rust proofing,protection of cutoff material, and snowplow countermeasures. These measures also
lengthen service life, not only for expansion devices but also for bridges themselves,
and result in higher LCC assessment. In this paper, an improved expansion device andits performance evaluation method were presented in consideration of the above
matters. In the future, the authors intend to summarize the experimental data obtained
from the wheel running test and the driving test on an actual bridge to confirm thetransmission mechanism of wheel loads.
Reference[1] The Plazwire System – a long-term anti-rusting technology involving plasma
spraying, Kyushu Electric Power Co., Inc. Research Laboratory

218

FATIGUE AND CORROSION IN CONCRETE DECKSWITH ASPHALT SURFACING
Yoshiki Tanaka1
Jun Murakoshi1
Yuko Nagaya1

Abstract
In 20th century, a large number of highway bridge concrete decks in U.S. suffered fromcorrosion due to deicing salt, but those in Japan suffered from fatigue due to cyclic loadingof heavy trucks. Lately, significant corrosion of concrete decks due to deicing salt havebeen reported in Japan, giving rise to further concerns about combined deterioration fromfatigue and corrosion in existing bridge decks designed according to old specifications.
This paper provides chloride profiles in concrete decks surfaced with asphalt mixtures,
which is one of reasons concerning a significant difference in deterioration between bothcountries. Additionally, with regard to mechanisms of fatigue deterioration of concrete
decks, a current result obtained from the truck wheel traveling tests is presented.

Introduction
In 1960s, construction of interstates began in U.S. But spalling due to corrosion frequently
appeared on concrete decks in several years since completion. It was caused by a rapid
increase in salt use for deicing in those days shown in Fig. 1. In 1970s, the deterioration
affects traffic safety, drivability, traffic congestion during repair, and budget, being a major
problem. A large number of deficient bridges due to corrosion still remain as current issues
in bridge management in U.S.1
In Japan, fatigue due to cyclic loading of heavy trucks is major deterioration of
concrete decks shown in Fig. 2 to date. Currently, however, attention should be paid not toonly fatigue, but also chloride-induced deterioration on the concrete decks, because salt
use have been increasing since spike tires were prohibited in 1990s as shown in Fig. 3.2, 3
Figure 4 shows chloride-induced deterioration observed on a bridge deck in Japan the late
1990s. While corrosion due to deicing salt is not often observed in national roads and ruralroads, significant corrosion on bridge decks in expressway was reported.4 In addition, mostof existing bridges were built from the mid-1950s to the mid-1970s, having lower potential
for fatigue durability than current bridges (Fig. 5, 6). The possibility of combined
deterioration from fatigue and corrosion should be taken into account in maintenance ofexisting highway bridges.
1

Public Works Research Institute, Japan219

With respect to the reasons why both countries differ in major deterioration of
concrete decks, quantity of salt use, traffic loading including illegal vehicles and asphalt
surfacing (newly constructed bridges are almost surfaced with asphalt pavement in Japan)
contrast apparently (Fig. 7).2, 3 Chloride ingress from the asphalt surface to concrete decks
is probably governed by water. As-built asphalt mixtures should not be expected aswaterproofing because of its permeability, whereas it was reported that asphalt surfacing
contributed to prevent concrete decks from salt ingress while being imperfect.3
Additionally water significantly affects fatigue durability of concrete decks.5 Thus field
performance of asphalt surfacing on the decks should be investigated for managing toprevent existing concrete decks from fatigue and corrosion in Japan.
This paper indicates results of field investigations on asphalt surfacing and chloride
profiles in decks using two closed bridges; a post-tensioned concrete bridge at coastal area
(ATG bridge) and a steel girder bridge with a concrete deck (TYD bridge) subjected todeicing salt. In addition, new results regarding mechanisms of fatigue deterioration in
concrete decks are presented for discussing the influence of spalling on fatigue behavior.
Field Investigations of Chloride Ingress under Asphalt Surfacing
Test cores consisting of asphalt mixtures and portland cement concrete of decks were takenfrom the closed bridges before removing asphalt surfacing (Fig. 8). Content of acid-soluble
chloride ions in each slice with 10 mm depth of concrete cores was determined. A
percentage of void content of asphalt mixture cores Vv was determined as described below.
Vv = 1 - (γM / γmax) × 100%
(1)where, γM is measured mixture density (= dry weight in air /(wet weight in air - weight inwater)); γmax is a percentage of the maximum density. The values of γmax were determinedusing a fragment of the removed asphalt mixtures according to the test method similar to
ASTM D 2041.
Water permeability tests of asphalt mixture cores were conducted under a constantpressure of 150 kPa. Water permeability k was determined from the test results based on
Darcy's law as below.
k = LQ / (AHT)(2)
where, Q is total discharge (cm3); H is head (cm, =1 500 cm); T is time (s); A is area ofcores (cm2); L is depth of cores (cm). No leakage of water in 24 hours was defined as
impermeable.
In Fig. 9, chloride profiles of cores taken from ATG bridge are shown. The ATG
bridge was built at coastal area in 1971, replaced due to corrosion of post-tensioned girders
in 2005. No corrosion was observed on the road surface. Curbs were replaced around 1985
for upgrading traffic safety. The thicknesses of asphalt mixtures in the cores ranged from
37 mm to 66 mm. From the figure, it can be seen that chloride contents at the surface ofdeck concrete were much less than that of a core A-3 taken from the top of curb notcovered with asphalt, even though low chloride permeability of prestressed concrete wastaken into account. All asphalt cores of ATG bridge were determined as impermeable.

220

Likewise, results from the investigation of TYD bridge are shown in Fig. 10. The
water permeability k is indicated, no indication of k means impermeable except a core T-9
untested. Deicing salt was applied on TYD bridge, being somewhat apart from the sea. The
thicknesses of asphalt mixtures in the cores ranged from 44 mm to 71 mm. Chloride ingressof deck concrete was definitely observed at a joint between asphalt mixtures (core T-9).
Except the joint, much chloride contents were not observed in the concrete of decks. Allasphalt cores taken from TYD bridge were not impermeable. Determined permeability
ranged from 1 to 20 × 10-7cm/s except cores near curbs.
Most of asphalt cores taken from existing bridges including 20 bridges investigatedpreviously3 and the above mentioned ATG and TYD bridges had the void contents Vv ofless than 2%, being determined as impermeable. It was reported that the void contents ofasphalt mixtures decreased with increasing the service time.6 With regard to the decreasein the void content in service, compaction due to traffic loading probably influences. But
since impermeable areas were not observed only in loading areas determined by ruts, there
must be reasons other than the effect of compacting in service, such as melting due to hightemperature in summer and clogging. Although waterproofing was installed on the TYD
bridge since 1996, a fact that deicing salt had been applied since the day before theinstallation indicates that the decks was not prevented from chloride ingress by only
waterproofing.
Several routes of chloride-contaminated water from the road surface to decks seemto exist depending on depth of asphalt mixtures, magnitude of compaction, finishing at
joints of surfacing, damage level and maintenance level of surfacing (Fig. 11). After
chloride-contaminated water reach on the concrete decks, chloride ions penetrate into
concrete. Chloride penetration in uncracked concrete is ordinarily expressed by Eq. (3). 7




x 
C ( x, t )  C o 1  erf
2 Dc t (3)where, x is depth from the surface; t is time; Co is chloride level at the surface of concrete;erf ( ) is error function, Dc is diffusion coefficient of chloride ions in concrete.
As shown in Fig. 12, the diffusion coefficient obtained from ponding tests in salt
water significantly differs from that obtained from exposure tests in airborne chloride
environment at coast.8 Consider the diffusion coefficient of top cover concrete in decks,
while the condition of bare concrete of curbs corresponds to that of the exposure tests, thecondition of concrete decks covered with chloride-contaminated water under asphalt
mixtures is probably close to that of the ponding tests indicating rapid chloride penetration.
In addition, water ingress accelerates disbonding between asphalt surfacing and concrete,
providing a fine soil layer in the boundary. The soil layer keeps the surface of concrete wet
for long time, forming the condition closer to the ponding tests.
When chloride ingress continues, chloride level at reinforcing steel bars reach thethreshold level initiating corrosion of steel in concrete. Expansion of the steel bars due to221corrosion yields cracking and spalling of concrete as well as asphalt mixtures.
Curve fitting to Eq. (3) using chloride profiles obtained from decks without asphalt
surfacing provides two parameters of the boundary chloride level Co and the diffusion
coefficient Dc. Then exposure time t is simply determined as the term from construction.
But the chloride profiles obtained from decks surfaced with asphalt mixtures are notavailable to determine the diffusion coefficient of the decks because the exposure timegoverned by quality, durability and maintenance level of asphalt surfacing is unknown.
As an attempt, the exposure time is estimated using the diffusion coefficient
predicted from Fig. 12. Since water-cement ratios of old structures are not exactly
predictable, average values of reinforced concrete and post-tensioned concrete are roughly
assumed as 0.55 and 0.43, respectively. Fig. 13 shows the estimated exposure time of theconcrete decks under asphalt surfacing. With regard to the results of wet condition, theestimated exposure time of decks on ATG bridge is only 0.2 to 1 year despite the severe
chloride environment at coast and the lifetime of 34 years. In cores taken from TYD bridge,
obvious chloride profile was obtained from only core T-9 at a joint of asphalt, while
deicing salt was applied for long years. The estimated exposure time at T-9 is 8 years. It
cannot be determined whether chloride ingress took place after installing a waterproof
layer or before.
Consequently, it was found that although both bridges investigated did not had
waterproofing at completion, the expected exposure time was much shorter than theservice term. The results indicate that the impermeable or low permeable asphalt mixturescontributed to prevent bridge decks from chloride contamination in a manner.
Mechanisms of Fatigue in Reinforced Concrete Decks under Traffic Loading

Concrete spalling due to corrosion on decks influences fatigue durability of concrete decks.
Fatigue mechanisms of reinforced concrete decks subjected to cyclic wheel loads havebeen studied since 1960s. Some types of truck wheel traveling tests for real-size decks
were proposed to simulate real fatigue deterioration in 1980s,9 contributing various
research of bridge decks so far. To identify the influence of spalling on fatigue behavior,a new approach focusing on the change of internal stress with crack propagation was
attempted in our study using a specimen N for the test (Fig. 14).10, 11 From the results, atrelatively early cycles, arch action consisting of concrete arch rib and steel ties (lower mainsteel reinforcing bars) was recognized through a truck wheel traveling area, similarly to ashear resistance mechanism of reinforced concrete beams.
As shown in Fig. 15, cracks were observed in a deck specimen after the truck wheel
traveling test. For showing one of the behavior indicating arch action, distributions of axial
forces acting on a main tensile reinforcement at the center of the deck are shown in Fig. 16,
being measured using strain gages. The distribution at 50 cycles shows typical flexural
behavior of a slab. As the number of cycles increases, the distributions changed eventhrough the bar, indicating the behavior of the tie bar. The change was recognized at earlystage corresponding to the cycles Ns that an apparent neutral axis calculated using
measured strains of upper and lower steel bars turned to increase as shown in Fig. 17.

222

Similarly, using previous data of 11 specimens,12 the cycles to the change of theapparent neutral axis Ns were calculated. The cycles Ns corresponded to the cycles to thecrack density (crack length per area) reaching 90% of the final crack density, ranging from0.05 to 0.3 Nf,9 where Nf is the cycles to failure. All specimens were designed according tothe old specifications in 1964 except a specimen having details of 1972 specifications.
Fig. 18 shows the cycles Ns of the 11 specimens and the specimen N with theapplied load level normalized by Psx, where Psx is the reduced punching shear strength5
considering an equilibrium of forces at failure under cyclic loading. From the figure, it was
found that the cycles Ns related to developing arch action can also be expressed as a kindof fatigue behavior similarly to Nf.
Fatigue tests in compression using 100 mm dia. cylinders made of the same batchof concrete for the deck specimen N were conducted. A relationship between the measured
strain ranges at the initial cycle and the cycles to failure was obtained. From the result, themeasured strain ranges of 800 to 1000 microstrains in compression resulted in fatigue
failure at several hundred thousands cycles corresponding to the cycles to failure of 0.36
million cycles for the deck specimen N.
During the truck wheel traveling test of specimen N, after arch action was formed,
the regions, where the strain ranges measured using embedded gages in concrete exceeded
the strain range of 800 microstrains, gradually appeared. The deck specimen shifted into
the situation that concrete in the region could fail owing to fatigue in compression.
It is known that an elastic modulus of concrete decreases with cycles in fatigue tests.
To estimate the elastic modulus of concrete in the deck specimen N, each share of concreteand main compressive reinforcing bars in the internal compressive resultant force underarch action was calculated using measured strains of upper and lower reinforcing bars asshown in Fig. 19. Subsequently, assuming that the share equals to the ratio of concrete area
to steel area considering young's modulus ratio, the elastic modulus of concrete in theinternal compression zone of arch action was estimated. The result is shown in Fig. 20. It
can be recognized that the elastic modulus of concrete in the compression zone rapidly
decreased after developing arch action. The turn to increase of the apparent neutral axis
observed after arch action was caused by the increase of strains in the upper main
reinforcing bars due to the decrease of elastic modulus of concrete in the compression
zone.
From this viewpoint, spalling on decks definitely affects fatigue durability ofconcrete decks under heavy traffic conditions. In fact, another deck specimen having
artificial spalling on the top showed a significant decrease in the cycles to failure (Nf = 0.14
million cycles) when compared with the specimen N.11
Conclusions

From our investigations regarding fatigue and corrosion of concrete decks, the following
two topics were introduced.
1) In several surveys to date, asphalt surfacing often prevented concrete decks fromchlorides in a manner regardless of waterproofing. Most of asphalt mixtures covering223

old removed bridges were determined as impermeable or low permeable. Asphalt
mixtures seems to be getting impermeable relatively early in service, while being
imperfect.
2) At early stage during the truck wheel traveling tests of full-scale concrete decks, arch
action consisting of concrete arch rib and steel ties similarly to a shear resistancemechanism in reinforced concrete beams was recognized through the truck wheel
traveling area. Unless arch action is constructed, concrete in decks is out of the range
available to be fatigued in compression (Fig. 21). The cycles to arch action are likely to
be expressed as a kind of fatigue behavior.
Acknowledgments

The authors thank the many staff members in the Ministry of Land, Infrastructure,
and Transport that assisted with the surveys presented in this paper.
References
1. Paul Virmani et al., Corrosion Cost and Preventive Strategies in the United States, Federal Highway
Administration, FHWA-RD-01-156, March 2002.
2. Report of Research Committee on Deterioration of Concrete Structures due to Deicing Salt, Japan
Concrete Institute, Tokyo, Japan, Nov. 1999. (in Japanese)
3. Public Works Research Institute, Investigation on Life Cycle Cost of Concrete Bridges-Deterioration and
Maintenance Cost of Concrete Bridge, Technical Memorandum of PWRI No. 3811, Tsukuba, Japan,
March 2001. (in Japanese)
4. Yokoyama, K., Honjo, K., Kuzume, K. and Fujiwara, N., Study on deterioration mechanism of reinforced
concrete slabs with steel corrosion in expressway bridges, Proc. of 6th Symp. on Decks of Highway
Bridges, JSCE, pp.145-150, June 2008. (in Japanese)
5. Matsui, S., Fatigue strength of reinforced concrete slabs of highway bridges by wheel running machine andinfluence of water on fatigue, Proc. of JCI, 9-2, pp.627-632, 1987. (in Japanese)
6. Q & A on Pavement Technology, Kensetsu-Tosho, Tokyo, Japan, pp.77-82, 1972 and pp.179-183, 1983. (inJapanese)
7. Sherman, M. R., McDonald, D. B. and Pfeifer, D. W., Durability Aspects of Precast, Prestressed Concrete-Part 1: Historical Review, PCI Journal, V. 41, No. 4, pp. 62-74, July-August, 1996.
8. Tanaka, Y., Fujita, M., Cheong, H., Watanabe, H. and Kawano, H., Chloride Permeability of
High-Strength Concrete, Proceedings of fib 2002 Osaka Congress, pp. 145-154, October 2002.
9. Matsui, S., Study on fatigue of concrete slabs for highway bridges and its design method, Thesis, Osaka
Univ., Japan, Nov. 1984. (in Japanese)
10. Nagaya, Y., Murakoshi, J. and Tanaka, Y., Behavior of cracks in reinforced concrete decks under cyclic
running wheel loads, Proc. of JCI, 30-3, pp.907-912, July 2008. (in Japanese)
11. Tanaka, Y., Murakoshi, J. and Nagaya, Y., Influence of Spalling of Upper Cover Concrete on Fatigue
Behavior of Concrete Bridge Decks, Proc. of JCI, 30-3, pp.913-918, July 2008. (in Japanese)
12. Experimental study on the fatigue durability of highway bridge slabs, Technical note of NILIM, No.28,
Tsukuba, Japan, March 2002. (in Japanese)
13. Highway Deicing - Comparing Salt and Calcium Magnesium Acetate, SR235, TRB, National ResearchCouncil, 1991.
14. Bridge inventory data was provided by FHWA in 1999.
15. Shimomura, C., Sakai, Y. and Nakajima, H., Investigation on the real condition of spreading deicer, Civil
Engineering Journal, 31-5, pp.280-285, May 1989. (in Japanese)

224

12

Corrosion
From 50s to 60s- Increase in traffic volume- From sand to salt
- Improving equipments

Salt use (×106 tons)

10
8
6
4
2
0

50 52 54 56 58 60 62 64 66 68 70 72 74 76 78 80 82 84 86 88

Year

Fig. 1 Trends in deicing salt use in U.S. 13(a) Grid cracks(b) Through holeFig. 2 Deterioration due to fatigue in reinforced concrete decks

20,00020

Spike tires were prohibited.

凍結抑制剤散布量（ｔ）

Salt use (×1000 tons)

25,00015,000

TohokuregionProclamation

10
10,0005,000

00
昭和５６ 57

81

58

59

60 61 62 63 平成元年90
2
3
85

年度

4

5

6

7
8
95

9

10

11
99

Year

Fig. 3 Trends in deicing salt use in Tohoku region in Japan225

Fig. 4 Corrosion of top mat reinforcement due to salt use in reinforced concrete decks
Both photo shows top of decks after asphalt and spalled concrete cover were removed.
1
Level of deficiency
None
Some
Significant
Repaired

4000

0.8
Ratio

The number of bridges

6000

0.6

Level of deficiency
None
Some
Significant
Repaired

0.4

2000
0.2
0

0
26 39 56 64 67 72 73 78 80 90<0.25 0.5

1

2

3

5

10

10<

Truck volume (×1000 vehicles/12h）

Revised year of bridge spec.

Fig. 5 The number of steel girder bridges
classified under the level of deficiency due tofatigue in concrete decks (investigated in
1991)

Void content
and water
permeabilityof asphalt
mixtureFig. 6 The influence of traffic volume oflarge vehicles on fatigue in concrete decks
The ratio is based on the number of steel girder bridges
designed according to highway spec. in 1964.

(investigated in 1991)

Chloride
profile fromtop surface

Fig. 8 A core with 70 mm dia. taken from the road surface of a closed prestressed
concrete bridge at coastal area and purposes of the coreNote: Fig. 7 is shown in next page.

226

Comparisons of Deck Deterioration
between U.S. and Japan

Details of Decks
Bridge Deck Surface■ U.S.

■Chloride Attack■Fatigue

U.S.
Japan

Wearing Surface
Concrete
Epoxy
Bituminous
TimberGravelOther, None, N/A■ JapanAsphalt Pavement% of Deck Area
70.78%
4.87%
3.47%
1.09%
1.18%
16.89%
Almost(a) Major problems in concrete bridge decks(d) Comparison in surfacing14

Items of Comparisons

Quantity of Salt Use■Details of Decks■Materials - Concrete■Quantity of Salt
■Over-loading■ Massachusetts
19■ New Hampshire16■ Vermont17■ New York17
■ Michigan
13■ California
3(on State Highways, 1980s?)■ National Highway
0.3■ Expressway
2(Average 1982-1987)

Similar?
U.S. > Japan
U.S. < Japan

tons/lane-mile

tons/lane-mile(b) Key items of comparisons(e) Comparison in salt use13,15

Details of Decks
Deck Depth & Cover DepthOver-loading■ U.S.
Before corrosion

46 - 60s
70s
Late 70s

Deck Depth
6 in.
7-8 in.

■ U.S.

Cover Depth
1.5 in. (to center)
1.5 in. (to surface)2 in.

- 1.2-1.3 times of Design Loads■ Japan■ JapanBefore fatigue- 2 - 3 times of Design Loads(Investigated in 1984-90)

56 5.5 in.
1.2 in.
68 6.5 in. *1.2 in.
71 7.3 in. *1.2 in.
*) at Deck Span of 8.2 ft (2.5m)(c) Comparisons in details of decks(f) Comparison in overloading

Fig. 7 Comparisons in deterioration of bridge decks between U.S. and Japan(Presented by Y. Tanaka in a breakout session in Reno, 2000)

227

8

Curb A-3

3

Cl (kg/m )

6

4

4

C=53mm3

Cl (kg/m )

A-9

2

2

0

0
0

20
Depth (mm)

0

40

20

40
60
Depth (mm)

80

100

Sea side4

3

C=30mm2

Cl (kg/m )3

Cl (kg/m )

A-4
2

3

Drainage

A-10

Cl (kg/m )

A-1
4

4

2

0
0
0

40

4

3

A-2

2

3

2
0
0

20
Depth (mm)020Depth (mm)

2

40
0
0

4

40

A-7

20
Depth (mm)403

Cl (kg/m )

0

40

A-6

A-11
Cl (kg/m )20
Depth (mm)20
Depth (mm)4

Cl (kg/m )

4

40

3

20
Depth (mm)Cl (kg/m )00Lines of rut

0

4
3

Cl (kg/m )

A-12
2

2

Note:
Depth : depth from deck surface
C : cover depth
Unit of dimension : mm0020
Depth (mm)C=39mm400020Depth (mm)40Land side

Fig. 9 Chloride profiles in concrete under asphalt surfacing in a post-tensioned concrete
bridge exposed at coastal area for 34 years (ATG bridge)
4
Cl (kg/m3)

T-9
2

0
0

20
Depth (mm)Note:k : water permeability of asphalt (cm/s)
No data indicates impermeable
except T-9 at joint of asphalt pavement
was not measured.

40

4
Cl (kg/m3)

T-8
2

0
0

20
Depth (mm)404
Cl (kg/m3)

T-15
2

Drainage
0
40

2

0
20
Depth (mm)40

k = 1.1×10-7020Depth (mm)

4

40

2

0

T-14

2

0
0

20
Depth (mm)

4

T-11

3

2

0
0

4
T-12

Cl (kg/m )Cl (kg/m3)

Cl (kg/m3)

4

T-6

T-13
Cl (kg/m3)

4

T-7

3

4

Cl (kg/m )k = 2.1×10-6320Depth (mm)Cl (kg/m )040

k = 2.5×10-4

2

0
0

20
Depth (mm)40

k = 1.7×10-6

2

0
0

20
Depth (mm)

40

0

20
Depth (mm)40

Fig. 10 Chloride profiles in concrete under asphalt surfacing in a concrete deck of a steel
girder bridge used for 43 years suffering from deicing salt (TYD bridge)

228

Fig. 11 Routes of chloride ingress to decks surfaced with asphalt mixtures2

Diffusion coefficient Dc (m /s)

1.E-06
10-10
-11

10
1.E-07

Dc = 7×10-11e-1.1(C/W)
Ponding tests

1.E-0810-12
Dc = 5×10-11e-1.6(C/W)
Exposure tests-1310
1.E-0910-14
1.E-10
0.2

0.3
0.4
0.5
Water-cement ratio

0.6

Fig. 12 Difference in diffusion coefficients of chloride ions in concretes
between exposure test at coastal area and ponding test in salt water8
Estimated term of chloride
contamination (years)

100
ATG br. = 34 yrs.

10

Wet
Dry

TYD br. = 43 yrs.
Service term

1

0.1
A-1

A-4

A-6

A-7

A-9 A-10 A-11 A-12 T-9

Core #

Fig. 13 The estimated terms of chloride contamination at top of decks229

Fig. 14 Truck wheel traveling test using real-size concrete decks with 2.8 m x 4.5 m

Wheel load

Transverse

Fig. 15 Cracks in a deck specimen due to cyclic truck wheel loading

Tensile force of steel bar T (kN)

(Nf = 0.36 million cycles, wheel load = 157 kN)

50

Tensile forces were getting eventhrough the bar with cycles.

40

Number of cycles
N = 300 000 cycles30
N = 50 000 cycles

20

From slab action toarch action

Flexural behavior
was observed
N = 50 cycles
at early cycles.

10
0
0

250

500

750

1000

1250

Distance from the traveling axis x (mm)Fig. 16 Distributions of tensile axial force acting on a main reinforcing bar

230

1.0
Cycles to failure N f

Ns

0.7

20

0.5

40

P / P sx

Depth of neutral axis (mm)060CenterFailed area, 600mm from center

80100
1.E+0110

1.E+02102

1.E+03103

1.E+04104

1.E+05105

1.E+06106

Cycles to arch action N s
0.3
Spec. in 64
Spec. in 72

0.2

Plotted data indicates onlyN s of each specimen.
0.1
10

2

10

10

3

10

4

10

5

6

10

Number of cycles

Number of cycles N s or N f

Fig. 17 Relation of depth of the apparent neutral
axis from the top surface with cycles

Fig. 18 Relation of loading levelwith cycles to arch action

Fig. 19 Schematics of contributions of concrete and a top main steel bar
subjected to the internal compressive force under arch action

231

7

10

Estimated elastic modulus of concrete2at the compression zone (kN/mm )30

Measured elastic modulus of concrete cores : 26.5 kN/mm

2

20

10

0
10
1.E+01

Invalid estimation appeared prior
to arch action. The invalid results
are plotted to determine the pointof change to arch action.

102
1.E+02103
104
1.E+03
1.E+04
Number of cycles

105
1.E+05106
1.E+06

Fig. 20 Decrease in elastic modulus of concrete due to fatigue under arch actionB) Fatigue of concrete can beobserved under severe cyclic
compressive force under arch
action.
Center of span
Transverse direction
Top main steel bar

Wheel load:

Axis of arch actionjd

M=jd.T= jd.C:

Bottom main steel bar

xA) Arch action was recognized at early cycles. The internal
compressive force acting on concrete prior to arch action wasminer level for fatigue.

Fig. 21 Arch action and fatigue of concrete in compression

232

C

T

REPLACING THE SUSPENDER ROPES OF A TIED ARCH BRIDGE USINGSUSPENSION BRIDGE METHODSBARNEY T. MARTIN, JR., Ph.D., P.E.1, AND BLAISE A. BLABAC, P.E.2
ABSTRACT:The Thaddeus Kosciuszko Bridge consists of twin through-arch structures
which span the Mohawk River in the state of New York between the towns of Colonie
in Albany County and Halfmoon in Saratoga County. Working with Modjeski and
Masters, the contractor for this project proposed an alternative method for the
replacement of all 168 suspenders for these bridges utilizing a modified techniquepreviously applied to suspension bridges. One of the most significant benefits of the
proposed method was that it allowed the work to be performed under full traffic. This
alternative method saved the bridge owner approximately $5 million by eliminating
the complex jacking system and traffic control requirements shown on the originalContract Plans.1
INTRODUCTION
The Thaddeus Kosciuszko Bridge consists of twin through-arch structures
located in the State of New York which span the Mohawk River between the towns ofColonie in Albany County (on the South) and Halfmoon in Saratoga County (on the
North). The bridges were built in 1959 and each carries three lanes of traffic in asingle direction on the Northway (Interstate 87) – one bridge is for Northbound traffic
and the other for Southbound. The bridges carry a high traffic volume of
approximately 110,000 vehicles per day.

Figure 1 – Northway Twin Arch Bridges (Thaddeus Kosciuszko Bridge), Albany,
NY
1
2

President, Modjeski and Masters, Inc., Poughkeepsie, New York
Senior Design Engineer, Modjeski and Masters, Inc., Poughkeepsie, New York233

In February 2008, a contract was awarded for the replacement of all 168
suspenders on the bridges. The nearly 50-year old suspenders, consisting of 50 mm
diameter structural strands, had significant section losses as a result of corrosion
caused by an aggressive upstate New York environment. The winning bid pricesubmitted by Piasecki Steel Construction of just under $12 Million was over $4
Million less than the Engineer’s Estimate. A Value Engineering Proposal submitted
by the contractor subsequent to the award saved an additional $1 Million. Thus, thecontractor’s alternative method resulted in a combined savings of over $5 Million.
The cost savings were the result of an innovative approach to replacing thesuspenders using an alternative method from that shown on the contract plans. Thismethod substituted the complex jacking and shoring system shown on the plans with asimplified system similar to that used previously on suspender rope replacement
projects for suspension bridges. Using this system, the contractor was able to remove
and replace the suspenders under full live load. This was a significant improvement
over the method shown on the contract plans which required closing the bridge duringcertain phases of the work resulting in a complex and costly traffic plan that involved
a median cross-over to reroute traffic to the adjacent bridge.
BRIDGE DESCRIPTIONThe Northway Twin Arch bridges are two-hinged steel arches with a span of
182.9 m between the bearing pins and a rise of 36.6 m relative to the spring line.
Each bridge consists of two parallel rectangular arch ribs having a rectangular box
cross-section approximately 1.2 m wide by 2.4 m tall, built up of riveted steel plates
spaced 15 m center to center. Each bridge carries three 3.6 m lanes with two 1.0 m
shoulders, for a total deck width of 12.8 m. The suspended portion of the span has alength of 167.6 m with three approach spans on each end for an overall length of
232.7 m per bridge.
The deck superstructure is suspended from the arches with 50 mm diameter
structural strands – one pair of strands supports each end of the intermediate
floorbeams. The superstructure of each bridge consists of 21 intermediate floorbeams
spaced at 7.62 m on center (22 equal panels). The end floorbeams are connected either
directly to the arch (at the fixed end) or attached with a pinned link (at the expansion
end). Longitudinal stringers framing into the floorbeams support a reinforced concrete
deck (designed as non-composite, although shear connectors were provided for added
rigidity). Open steel grid deck is provided between the floorbeams in the shoulder
areas to provide drainage.
There are a total of 84 suspenders – the longest of which is over 29.87 m long.
The typical suspender has a Type 8 socket on each end – at the top, the socket bears
on a transverse beam within the arch box (see Figure 5); and at the bottom, it bears ona stiffened seat angle attached to the web of the floorbeam. The connection issymmetrical about the webs of the transverse beam and floorbeam.
The suspender connection at the crown of the arch is unique. At this location,
both suspenders share one large socket which in turn is supported by a pair of U-boltsthat straddle a large pin (see Figure 8). The pin is a vestige of an original design234

feature of the bridge – it originally functioned as the crown pin for a three-hinged
arch. As part of the original design, the bridge was constructed as a three-hinged archfor dead load. Once the deck superstructure was completed, the crown was locked byinstalling cover plates on the top and bottom of the box section to make this part of
the arch continuous. Thus, the bridge was transformed into a two-hinged arch for live
loads. Unfortunately, the design made no provision for future replacement of thesuspender – the splice plates effectively blocked access to the suspender connection.
This made for a more complicated repair at this location, as will be discussed later inthis paper.
PROJECT BACKGROUND
In recent years, the condition of the suspender strands had rapidly degraded
due to a combination of age and exposure. Salt spray from the adjacent roadway had
run down the suspenders over the years, concentrating the damage at the area where
the suspender connects to the floorbeam. The suspender strand in the vicinity of theseat angle began to suffer from corrosion related section loss, eventually leading to
failure of the individual wires. As the wires began to break, they would splay
outward, eventually resulting in a rather dramatic display of deterioration (see Figure3). In the most extreme cases, section losses of up to 50% were reported. As a result,the decision was made to undertake an immediate replacement of all 168 suspenders
on both bridges. Of these, a total of 22 suspenders were identified as requiring
emergency replacement.

Figure 2 – View at Crown of Arch

Figure 3 – Photo of Deteriorated
Suspender Strand with Broken Wires(After Removal

235

The contract, awarded in February 2008, required the replacement of the 22
worst suspenders prior to the end of November 2008 and complete replacement of allsuspenders by November 2009. The contract contained a penalty clause of $10,000
per day if the contractor did not meet the November 30, 2008 deadline.
Although a temporary support system for the replacement of the suspenders
was fully designed and detailed on the contract plans, the contract documents allowed
for an alternative lifting scheme to be submitted by the contractor. The contract
stipulated that any alternative system would have to meet the specific requirements ofthe contract, such as load capacity and deflection criteria, and had to be designed anddetailed by a licensed Professional Engineer. Furthermore, the alternate system would
be subject to review and approval by the NYSDOT.
The winning contractor, Piasecki Steel, hired Modjeski and Masters to prepare
the detailed design calculations and working drawings of the alternate structural
lifting system. The design was completed in May and accepted by the NYSDOT in
June 2008.
The contractor began replacing suspenders in September 2008. In October, asa result of the routine annual inspection of the bridge performed by the NYSDOT,
some of the 22 yellow-flagged suspenders were upgraded to a red-flag condition –thus requiring repairs to be completed within 6 weeks. This forced a modification of
the already advanced replacement sequence; however, the contractor was able tomobilize additional jacking systems to accommodate the accelerated schedule and therepairs were completed on time – with just 1 day to spare before the November 30th
deadline!
DESCRIPTION OF THE ALTERNATIVE STRUCTURAL LIFTING SYSTEM
The structural lifting system developed for this project is very similar to those thathave been used for the replacement of suspender ropes on suspension bridges. For
these structures, the lifting system is designed to perform the following tasks:
1. Temporarily support the floorbeam to allow the suspender to be removed
2. De-tension the existing suspenders
3. Tension the new suspenders
Although similar in concept, some of the features of the suspender system for thisbridge structure were quite different from the details found on typical suspension
bridges. Due to the unique features of this structure, developing a structural lifting
system to allow the completion of all the required repairs presented quite a challenge.
Some of these features include:Each pair of suspenders supporting the floorbeams of the arch consists of two
individual elements, anchored top and bottom. In this case, each suspender could
be removed and replaced individually. On most suspension bridges, thesuspenders are continuous over the top of the cable, such that both legs of thesame rope must be replaced simultaneously. Therefore, when replacing a236

suspender rope, both legs of the suspender are typically de-tensioned andtensioned simultaneously.
The suspenders of this bridge are attached directly to the web of the transverse
floorbeams. Unlike suspension bridges, this bridge has no longitudinal stiffening
element (truss or girder) that supports the floorbeams. On a suspension bridge, thepresence of the stiffening truss typically allows for the installation of separate
systems during rope replacement: one to temporarily support the panel point andanother to de-tension/tension the suspender ropes. In this case, the structural
lifting point had to be located directly on the floorbeam itself and allowed for little
space for the attachment of a separate system for suspender removal andinstallation. Furthermore, the floorbeams required steel repairs, some of which
could only be done once the floorbeam was off-loaded and the suspenders were
removed.



Due to the lack of a longitudinal stiffening element, the structure does not provide
for a redundant load path to redistribute loads to adjacent panels during theremoval of a suspender. Owing to the resulting low inherent stiffness, the ends ofthe floorbeams are relatively free to deflect during jacking operations. Therefore,
the original design contract specified strict limits on the deflection of thefloorbeam during the removal and replacement of the suspender: no more than 8
mm of differential downward deflection between any two adjacent floorbeams;
and no more than 6 mm of upward deflection of any one floorbeam from its
original position. The reasoning provided by the design engineer for the deflection
limits was to prevent cracking of the concrete deck as a result of excessive
movements of the floorbeam. However, due to the lack of longitudinal stiffness ofthe deck superstructure, the displacement of the floorbeam as a result of live load
exceeds that allowed during jacking (particularly near mid-span where the elastic
stretch of the suspenders under the design live load is on the order of 25 mm).

ADVANTAGES OF THE ALTERNATIVE SYSTEM
The method in the value engineering proposal has a number of advantages overthe system shown on the original contract plans, such as:


The system shown on the contract plans relied on transferring load to the adjacent
floorbeams (and deteriorated suspenders) with a temporary support beam (seeFigure 4). The alternative method uses a completely independent system tosupport the floorbeams during suspender rope replacement (see Figure 5).

With the alternative method, there is no change in load path during the removal
and replacement of the suspender. The jacking system is attached to both the archand the floorbeam in a way that is structurally equivalent to the existing
suspenders (see Figures 6 and 7).



The alternative structural lifting system does not require modification of theexisting arch cross-section to allow tensioning of the new suspenders. Tensioningof the new strands is accomplished by the same friction clamping system used to
de-tension the strands. The stressing method shown on the original contract plans

237

involved cutting a permanent hole in one of the plates comprising the rib crosssection in order to tension the new suspender from within the arch using a special
type of top socket.

Figure 4 – Method of Replacing Suspenders in Original Contract Plans.


The alternative method allows for suspender rope replacement under full live load.
Furthermore, the contractor fabricated the new ropes using the measurements
shown on the shop drawings from the original bridge construction, rather thanattempting to measure the suspenders in place. This eliminated the complex andcostly traffic control procedures required by the contract plans that involved
shutting down one bridge at a time and installing cross-over’s to re-route traffic to
the adjacent bridge.



The alternative system combines all the functions of the temporary support, detensioning of the existing suspenders and tensioning of the new suspenders in one
system. This efficient system has a total weight of only 1.8 tonnes. The system
shown on the contract plans utilized a separate means of temporary support,
consisting of either beams supported by adjacent suspenders (for intermediate
panels) or towers supported at ground level (at end panels). The temporary
support beams alone consisted of double W36x300’s with a weight ofapproximately 15.5 tonnes. The temporary support towers had a height ofapproximately 7.3 m, a capacity of 159 tonnes and required a temporary footing at
the edge of the river.



The contractor was able to complete the replacement of all the suspenders byfabricating 4 identical systems.

238

Figure 6 – Connection at Transverse
Beam Inside Arch Box

Figure 7 – Structural Lift System
Installed at Panel Point 3

Figure 5 – Alternate Structural
Lift System

239

FEATURES OF THE ALTERNATIVE SYSTEM
A thorough description of the features and functions of each element of thealternative jacking system is beyond the length limits of this paper. The following
section highlights the details of some of the more unique elements of the alternative
system; these are: 1) the friction clamp, 2) the floorbeam steel repairs, and 3) the steel
repairs at the arch crown (Panel Point 12).
FRICTION CLAMP – Rope removal and installation was performed using afriction clamp, similar to those used on previous suspension bridge suspender rope
replacement projects. According to the contract specifications, the clamp had to be load
tested to determine the friction factor of the suspender strands for use in the design of theclamp. Due to the proposal to use the friction clamp for the installation of the new ropes,
the load test was also required to demonstrate that there would be no damage to thegalvanized coating of the suspenders as a result of the applied clamping force.
For this project, it was elected not to use a zinc liner at the contact surface
between the clamp and the suspender. Previous experience with installing clamps on wirerope suspenders indicated that there would be no damage to the surface of the ropes. The
contractor pointed out that clamping devices are used during the manufacturing of bothropes and strands (as part of the stressing beds used to pretension the strands to remove
the initial elastic stretch) which do not utilize zinc liners.
The contractor fabricated two friction clamps for the load test based on apreliminary design performed by Modjeski and Masters (see Figure 8). Assuming afriction coefficient of 0.15, the clamps were designed for the required structural lifting
load of 445 kN (per suspender) with a safety factor of 1.5 against slip (ultimate capacityof 668 kN). This resulted in a clamp utilizing 20 – 25 mm diameter A325 bolts fully
tensioned to a minimum load of 227 kN each (total clamping force of 4,536 kN, min.).
For the initial test, the ram was operated to its maximum capacity (890 kN), but no slip
occurred in either clamp. This proved the design was adequate; however, in order todetermine the coefficient of friction, it was necessary to determine the load at which slip
would occur. Therefore, the test was repeated, removing two bolts each time, until slip
was achieved. Slip finally occurred at a load of 890 kNs after eight bolts had been
removed (12 bolts remaining). Assuming an average bolt tension of 271kN per bolt (227kN min., 316 kN, max.) with a total clamping force of 3,256 kN, the actual coefficient offriction coefficient was calculated to be 0.27. This meant that the actual capacity of theclamp was 1,223 kN (ultimate) vs. the required capacity of 445 kN, resulting in a safety
factor of 2.75 for the design jacking load.

240

Figure 8 – Friction Clamp During Load Test
Upon the conclusion of the test, the clamps were opened and the rope wasvisually inspected by NYSDOT personnel. The load test demonstrated that the galvanizedcoating was in satisfactory condition – there was localized deformation of the coating due
to the high bearing stress on the surface of the outer wires, but the integrity of the coating
was not compromised.
FLOORBEAM STEEL REPAIRS – The original Contract Plans showed three
types of steel repairs for the floorbeams: Type 1 involved replacement of the end stiffener
angles and plate; Type 2 involved replacement of the vertical stiffener angles at thesuspender connection; and Type 3 combined both the Type 1 and Type 2 repairs plus thereplacement of an existing fill plate between the vertical stiffener angles and the web.
The contract was later modified to include Type 2 repairs at all locations (adding
approximately $700,000 to the cost of the contract). This was prudent given that the only
time these repairs could be completed was when the suspenders were removed. An
inspection of the existing stiffener angles also revealed that many were in poor condition
and the outstanding legs were actually bowed outward (possible due to crevice corrosion
at the top and bottom of the stiffener angles). Type 1 repairs could be performed
independent of the suspender removal because no off-loading was required.
As a result, it was imperative that the jacking system design accommodate theneed to perform steel repairs once the suspenders were removed. In order to accomplish
this, the floorbeam would have to be lifted away from the existing suspender connection.
However, the most logical location – the bottom flange – was blocked by the presence of
the wind chord. Due to the detailing of the wind chord, as well as the deteriorated
condition of this member, transferring the floorbeam reaction through this member was
not deemed a viable option. Therefore, the decision was made to provide a temporary
jacking beam that would effectively bridge over the wind chord by supporting thefloorbeam at two points – at the end stiffeners and at the first interior stiffener. (see241Figure 4) A support plate connected to the hanger rods would react against the jacking
beam through a simple rocker bearing to ensure equal distribution of the lifting loads toall four hangers. This system required that any Type 1 repairs be performed prior to theinstallation of the jacking system, as the beam utilized the end angles for support. In
cases where no Type 1 repairs were required, the structural capacity of the existing
riveted connection of the end angles was compared to the anticipated lifting loads. In fact,
it was determined that the rivets were not sufficient and that either the rivets would need
to be replaced with new high strength bolts or that additional bolts would need to be
added between the existing rivets (the latter option was preferred by the Contractor).
At locations with Type 3 repairs, an additional step was required in the repair
process as a result of utilizing the end angles for support of the jacking system. In orderto replace the existing fill plate, it needed to be cut into two sections to allow the new end
stiffener angles to remain in place while the remainder of the fill plate was removed
along with the vertical stiffener angles. This cut had to be performed delicately in order toprevent damaging the existing web. The Contractor employed a core drill to make aseries of adjacent holes along the cut line to remove the majority of the plate material.
Then a grinder was used to remove any remaining material between the holes and to
make a straight edge. A new narrow fill plate was installed along with the new end
stiffener angles. Later, when the vertical stiffener angles were removed, the remainder ofthe fill plate was removed and replaced.
SUSPENDER REPLACEMENT AND STEEL REPAIRS AT CROWN OF
ARCH (PANEL POINT 12) – The contract plans originally stipulated an extremely tightwindow for the contractor to remove and replace the suspender strands at Panel Point 12(crown of arch). The work consisted not only of the replacement of the strand itself, butalso involved substantial steel work – the bottom splice plates of the arch first had to be
cut to re-establish the original access opening and allow the existing socket to be
removed. Once the replacement of the existing suspenders and hardware was complete,
the splice plates needed to be reinforced by installing new cover plates. Cutting theaccess hole involved removing a substantial amount of area from the existing splice
plates which maintained the continuity of the arch for live loads. Therefore, the Contract
Plans contained an involved procedure that restricted much of the work to take place
while the two lanes adjacent to the arch could be closed – a condition that was only
allowed during a 6-hour overnight window. The Contractor indicated to the DOT that itwas not physically possible to perform all the required repairs in such a limited timeframe(not to mention the difficulty associated with attempting the work at night). Upon further
discussion with Modjeski and Mastersn, the restrictions were modified to allow thesuspenders to be removed and replaced during one - two-lane closure window and thesteel repairs to be performed during a subsequent two-lane closure window. Actual
cutting of the splice plates was allowed while maintaining only a single lane closure.
While these work restrictions were ostensibly designed to limit the loads on thearch to account for the reduced splice plate section, a careful review of the procedure on
the contract plans showed that the restrictions would not have this effect. In fact, the242

procedure indicated that during certain times, lane closures were not required to be
maintained at all while the access hole was opened. This fact was confirmed with the
NYSDOT and the consultant responsible for the design. Their analysis indicated that thearch would effectively revert back to a two-hinged arch while the repairs were performed
and would be adequate for the anticipated live loads. Once it was realized that theburdensome procedure shown on the Plans served no real purpose, it was abandoned infavor of a much simpler method of affecting the repairs.
A repair was developed that allowed all work to proceed without the need for laneclosures, except for final torqueing of the bolts on one-half of the splice connection(approximately one hour’s work, during which a two-lane closure would be required).
The jacking system was similar in concept to the other panels on the bridge in that an
independent hanger system was used to carry the floorbeam reaction while the existing
strands were removed and replaced. The difference at this location was the connection at
the arch due to the presence of the original pin in place of the typical transverse beam. In
this case, saddle blocks were fabricated that sat on top of the pin which accepted the allthread hanger rods and transferred the hanger reaction directly through bearing on the top
surface of the pin (see Figure 9).

Figure 9 – Cross-Section at Crown of Arch (Panel Point 12)243

The tight clearances necessitated some trimming of the edges of the access
opening to provide clearance for the hanger rods. The resulting minimum clearance
between the new top socket and the hanger rods in the final assembly was only 8 mm.
This did not leave much room for error in the installation of the approximately 1.41
tonnes assembly (top socket with a pair of suspenders and bottom sockets). Once the newsuspenders, socket and U-bolts were replaced, new cover plates were installed to close
the access opening. The cover plates were installed with the bolts torqued on only one
side of the splice – final torqueing of the remaining bolts was only performed under a twolane closure to minimize the live load stresses on the arch, which are transferred throughthe splice plates (see Figure 10).

Figure 10 –Access Opening Repairs at Crown of Arch (Panel Point 12)244

CONCLUSION
A modified method typically used in suspension bridges was proposed as analternative method for the replacement of the suspenders for twin through-arch bridges
which saved the owner, the NYSDOT, over $5 Million on a roughly $16 Million project.
The alternative structural lifting system was not only less expensive and easier to installthan the version shown on the contract plans, but also safer in that it did not impose
additional loads on the existing deteriorated suspenders during the replacement process.
The system also improved safety and decreased inconvenience to the traveling public byeliminating nightly closures of the individual bridges, which required crossing all traffic
over to the adjacent bridge. Even though additional steel repairs have been added to theproject, the contractor is on schedule to complete the replacement of all 168 suspendersof both bridges by November 2009.

245

blankSUSPENSION BRIDGE CABLES: 200 YEARS OF EMPIRICISM, ANALYSISAND MANAGEMENT
Dr. Bojidar Yanev1
Abstract
Over the last 200 years suspension bridges have been at the forefront in all
aspects of structural engineering, including empirical and theoretical studies, andconstruction and lifecycle management. Their spans have grown from 50 to 2000 mwith designs for 3000 m under consideration. Key elements of modern suspension
bridges are their cables. The evolution of suspension cables is examined. The mostrecent developments in their design, maintenance and inspection are described.
Conclusions regarding suspension bridges and engineering structures in general are
drawn.
Introduction: Empiricism and Theory
Engineering structures must succeed in three domains: the theoretical, thephysical and the social. The operating tools of these domains are abstract analysis,
empirical application and economics. Whether successful or not, the required
synthesis is particularly spectacular in long-span bridges, among which thesuspension ones are the undisputed champions. Suspension bridges using natural fiber
ropes have existed since prehistoric times, but their modern history begins with iron
chains. The first United States patent for a bridge suspended on iron chains wasawarded to James Finley (1762 - 1839), a land-owning judge, in 1808. The patented
bridges had a stiffened roadway, designed to carry pedestrians and horse carriages
cost-competitively over spans up to 76 m. Judge Finley arrived at his modest but
reliable bridges by experiments. His "empirico-inductive" understanding of thesuspension structural scheme inspired him to forecast that "something further may bedone in the art of bridge building than has yet been accomplished" (Karnakis, 1997, p.
53). Finley appears to have been fully aware of the need for a satisfactory analyticalmodel.
Navier (1785 - 1836) formulated suspension bridge theory in his Memoire surles Ponts Suspendus in 1823. The analysis, design and construction of suspension
bridges in France is concisely and comprehensively reviewed in LCPC/SETRA, 1989.
In 1827 Navier's extensively analyzed 170 m span at Pont des Invalides, barely
completed, had to be dismantled, primarily due to the malfunctioning anchorages.
Ever since, analysis and experiment have relentlessly challenged and stimulated theart and science of bridge building. Marc Seguin (1786 - 1875) complained that Navier
treated "theoretical notions or mathematical solutions [as] a priori admissible"(Karnakis, 1997, p. 269). Nonetheless, further developments, including Seguin's, owemuch to the theoretical backing of Navier's Memoire.
______________________
1
Executive Director, Bridge Inspection & Management, New York City DOT,
Adjunct Professor, Columbia University247

In the American Railroad and Mechanics Magazine of April 1, 1841, No. 379,
Vol. XII, J. Roebling (1806-1869) argued that "the successful introduction of cablebridges into the United States would require the combination of "scientific knowledge
and practical judgment of the most eminent Engineers". Roebling proved singularlycapable of providing both, along with the organizational abilities required for
manufacturing the high strength galvanized cable wires and the procurement of thefinancial backing his monumental bridges needed. Thus his Brooklyn Bridge in New
York City (main span 487 m), completed in 1883 by his son and daughter in law,
Washington and Emily Roebling, is carrying more than 80,000 passengers daily in2009. Billington (1983) concluded that the integration of theory and practice achievedby Roebling and Eiffel (1832 - 1923) elevated structural engineering into an art. One
disadvantage of great artistic accomplishments, however is that they do not lendthemselves to standardization and mass production. Engineering practice, in contrast,
must deliver utility repeatedly, reliably and cost-competitively. This apparent
contradiction is demonstrable in bridge design where spans shorter than 150 m aresubject to detailed specifications by AASHTO (American Association of State
Highway and Transportation Officials), but the longer ones are not. Neither are cablesupported bridges of any length. Thus in the domain of the foremost structuralaccomplishments, failures and successes jointly advance the state of the art.
Failure: the Ultimate Test
Despite the advances in abstract analysis and controlled testing, failures havethe most conspicuous influence on bridge design, construction and management. No
planned engineering experiment can match the social and professional impact of anunintended structural collapse. The shock value of the suddenly perceived ignorance
and the usually considerable losses add up to a lesson easy to understand andremember by non-professionals. Bridge failures in the United States from 1966 to2005 are attributed to various respective causes in Fig. 1. Natural hazards (and mostly
hydraulic-related ones) dominate this listing as they do in most classifications.

Figure 1. 1502 bridge failures in the USA, 1966 – 2005 (courtesy STV Engineers)248

Suspension bridges represent a small portion of the total bridge population,
however they are essential to transportation, and at the forefront of engineering
innovation. Consequently their failures are critically important not only quantitatively
but also qualitatively. H. Petroski (1993) regards failures are inherent in the creative
process of bridge design and construction. Citing earlier work by Sibly and Walker inEngland, he argues that each innovative bridge form is developed by trial and error
until its limits are surpassed and spectacular failure occurs. Only then does theory
catch up with the practice and fully explains the structural behavior. By this estimate,
the still evolving cable-stayed bridges should be viewed with particular concern. A
few popular examples may add a historic perspective to this controversial subject.
19th Century cable-supported bridges
Nineteenth Century engineers experimented with both suspension and cablestay bridges using both ropes and chains. The more complex dynamic behavior ofthese structures resulted in greater lapses in the designer's knowledge, and hence, themany failures under wind loads, as well as those caused by rushing crowds.
The cable-stayed bridge at Dryburgh collapsed in a gale storm in 1818, 6
months after it was completed. Samuel Brown's Brighton Chain Pier was destroyed bya storm in 1836. The landmark bridge crossing the Menai Straights in England with amain span of 177 m was designed by Thomas Telford (1757 - 1854) and completed in
1826. The bridge chain suspension system never failed, but high winds caused
significant damage and repairs were necessary at least twice.
With a main span of 308 m, Charles Ellet's (1810 - 1862) Wheeling suspension
bridge over the Ohio River was the world's longest. Ellet had to repair it after it wasdestroyed by wind in 1854. Roebling later replaced it with his hybrid suspension/staysystem. That system created enduring dynamically stable structures without a rigorous
theoretical backing. Theoretical advances in the 1920s enhanced suspension bridge
analysis. Roebling’s semi-empirical hybrid system was stripped of the diagonal stays
and the stiffening trusses, ultimately producing the aerodynamically unstable Tacoma
bridge.
Tacoma Bridge (1940)
The bridge, in Washington State, was designed by the early proponent of large
displacement theory Leon Moiseiff (1872 – 1943), renowned for his contributions at
Manhattan Bridge in New York and the Golden Gate in San Francisco. Distinguishedby a relatively narrow cross section and longitudinal stiffening girders in lieu of the
traditional trusses, it collapsed under wind loads after months in service. The failure is
attributed to ignorance of the dynamic phenomenon later recognized as flutter. Yet
flutter had already been identified by Von Karman (1940) in research on aircraft wingstability. Thus the ignorance could be perceived as design error. The new Tacoma
suspension bridge has a deep stiffening truss. In 1937 Othmar Ammann (1879 – 1966)
won an award for the design of the Whitestone Bridge in New York City. Following
the collapse of the Tacoma, the very similar Whitestone was stabilized with stiffening249

trusses, diagonal stays and a tuned mass damper mid-span. None of theseenhancements proved conclusive. They have been removed or are reassessed during acurrent structural reconfiguration.
Silver Bridge (1967)
The bridge over the Ohio River at Point Pleasant collapsed due to a brittle
fracture of an eye-bar of the suspension system. Poor quality control during
construction, inability to inspect, ignorance of fatigue behavior and lack of
redundancy all contributed to a degree that is still debated. The variety of contributing
causes entailed similarly wide ranging responses. The non-redundant 2-eye-bar
suspension chain was entirely discredited, leading to the controlled demolition of asimilar structure over the Ohio River at St. Mary’s, West Virginia. Impressed by thecollapse, which took 46 lives, the United States Senate mandated biennial bridge
inspections, clearly recognizing that the lack thereof is a management deficiency. The
outcome was the National Bridge Inventory (NBI) which currently contains data on
more than 600,000 vehicular bridges, 200,000 railroad bridges and more than 100,000
culverts. Fracture-critical details became an important part of the inventory andspecial provisions were designed for their inspections.
Partial failures
Partial failures, particularly when they appearas near misses are easily
overlooked and their lessons are harder to learn. For example, a database developed at
Cambridge University blames deterioration for only 3% of the considered bridge
failures. Yet deterioration is recognized as widespread. The losses it has caused arevast but not quantifiable, because they rarely include fatalities. As a result,deterioration-related failures are on the rise. They can be traced to neglectedmaintenance, as well as to the design which did not anticipate such neglect. Since
deterioration is a relatively slow process (compared to natural hazards, fractures andlosses of stability for example), its effects have been intercepted by inspections andarrested by rehabilitations. Thus partial failures have become increasingly significant.
Both Pont de Tancarville and Pont d’Aquitaine in France had their suspension cables
replaced without incident following the discovery of breaks caused by corrosion. The
original cables had no wrapping and consisted of helical non-galvanized strands. The
4 cables of the Williamsburg Bridge (1903, 488 m main span) in New York City were
rehabilitated in the 1990s after a long debate over their viability. The cables, eachconsisting of 7696 non-galvanized parallel 5 mm wires grouped in 37 strands, wererapidly corroding. During the rehabilitation several strands were re-anchored, manybroken wires were spliced, corrosion-inhibiting oil was added and water-proofing was
supplied in the form of new wrapping.
Many suspenders and stay-cables have broken, however the effect of thesefailures has been localized by redundancy. In 1981 one of the diagonal stays of the
Brooklyn Bridge broke due to corrosion and killed a pedestrian. All suspenders andstays were subsequently replaced. A suspender broke at the first Bosporus Bridge in2004 and a stay cable was burned by lightning at the Rion – Antirion Bridge in 2006,250

neither one causing significant traffic interruptions. Suspender and stay replacement isrecognized as a periodic necessity. The suspenders of Manhattan Bridge have been
replaced at least once and will be replaced again under a pending contract.
Overview of failures
Structural failures in general do not lend themselves to purely quantitative
assessments. To the inability to quantify loss of life and other user costs, there are
difficulties with evaluating the loss of public and professional confidence. Forensic
investigations are post-event efforts to eliminate failure causes. Once identified suchcauses are perceived as vulnerabilities requiring special treatment. The identificationand elimination of vulnerabilities is a perpetual task of management depending on
every specific circumstance, however some common characteristics have emerged(also discussed in Yanev, 2007). Failure causes have been classified according to avariety of criteria including the following:- Event-based
The classifications of the Cambridge database and Fig. 1, cited earlier are ofthis type, however they also refer to the structural type and material.
- By the mode of material non-performanceThis classification would identify metal fatigue and fracture, corrosion,
corrosion fatigue, ductile failure, residual stress, yield, shear, concrete fatigue,
chemical reactivity, temperature, and so on. Although helpful, this classification
misses certain vulnerabilities, such as instability. Thus another classification becomes
necessary, distinguishing between local and global behavior. Whereas a material
failure is by definition local and can lead to global consequences, instability is a
global failure which causes local material non-performance.
- "As designed" and "not as designed" modes (Thoft-Christensen and Baker, 1982)
The failure of the deck truss bridge carrying Rte. I-35-W in Minneapolis on
Aug. 1, 2007, caused by overstress of poorly dimensioned gusset plates, is in the “asdesigned” mode. These are practical failures resulting from poor execution. The
failure of the Tacoma bridge, caused by the hitherto unknown flutter is in the “not as
designed” mode. Such failures point o theoretical deficiencies.
- According to the bridge life-cycle, e.g. design, construction, maintenance, andoperation during which the failure occurs (Yanev, 2007)
This approach reveals that failures occur when each phase excessively relies on
the others, whereas it should have assumed responsibility as if it were critical. For
example after the collapse of the bridge over the Schoharie Creek (New York State,1987) caused by scour, it was concluded that the bridge could have survived if
underwater inspections had been more effective, however the collapse could have also
been prevented by selecting a longer span with piers out of the channel.
- No failure classification can be entirely independent, and neither are the revealed
vulnerabilities. A redundant approach to identifying vulnerabilities by various

251partially overlapping criteria has a better chance of capturing all possibilities. New
York State for example has established screening procedures for the following
vulnerabilities: hydraulic, seismic, overload, steel details, concrete details, collision,
sabotage.
- Significant failures are usually caused by combinations of two or more
vulnerabilities. In all of the preceding examples, design, construction andmaintenance might have intervened to avert the disaster, if they had seen their own
role as the most critical. The investigation following the collapse of the Sungsu truss
bridge in Seoul in 1994 concluded that the bridge had been "poorly designed, built,
maintained and used".
- Significant failures invariably exhibit a discontinuity in the process, in the product orin both. A common one is between experience and theory. This can be expressed as
an over-reliance of each on the other or as an ignorance of each about the other. For
example, the rationale justifying the use of non-galvanized high strength cable wires
on the Williamsburg Bridge was summarized as follows: “If the wires are maintained
dry, they do not need zinc coating. If not, it will not save them in the long term.” This
reasoning may appear sound, yet it lacks both the practical knowledge of life-cycle
cable performance and a theoretical model of wire deterioration. The opposite
solution would have been correct: wires should have been locally protected bygalvanization, whereas the cable should have been globally protected by superior
wrapping. As in all discontinuities, the remedy is redundancy.
- Failures can be traced to deficiencies in both the engineered product and the
managed process. Thus the gap between theory and practice is matched by an equally
detrimental gap between the increasingly diverging professions of engineering andmanagement. The current trend is to rely on engineering competence during thedesign and construction of the product (in this case the bridge) and to assign theprocess of its operation to a management, guided primarily by economic
considerations. The latter typically minimize initial costs, while shortening thestructural life-cycles and increasing the demand for future expenditures.
- Most simplistic and yet inevitable are the failure assessments according to theamount of damage and the responsibility. All structural failures are to some extent
attributable to management and this is reflected in the way society reacts to them. The
failure of the Silver Bridge was at the origin of the NBI and the biennial bridge
inspection program, the partial failure of the Williamsburg bridge led to the reestablishment of the Bridge Division at the New York City Department ofTransportation, the numerous failures caused by earthquakes in California and inJapan have influenced the design and construction of bridges worldwide, the collapseof the I-35-W bridge in Minneapolis has stimulated the use of non-destructive
structural monitoring techniques. Appropriate as these measures are, they remain
reactive, whereas the purpose of engineering management and design is to anticipate.
The history of suspension bridge cables offers examples of such anticipation.

252

Suspension bridge cables
The temporary closure, in 1988, and subsequent rehabilitation of theWilliamsburg Bridge attracted much attention to the condition of suspension cables.
(That project is currently concluding at a total cost of approximately $US 1 billion.)
The affected structures are among the oldest, largest and historically most significant
nationwide. The Cincinnati - Covington Bridge was opened to traffic in 1867, theBrooklyn Bridge - in 1883, Williamsburg - in 1903 and Manhattan - in 1908. Ten
major suspension bridges, including Ammann’s George Washington (1931 and 1957,
1068 m main span), Verrazano (1964, 1300 m main span), Triborough, Whitestone,
Throg’s Neck and the East River bridges are in the New York Metropolitan area.
During the 1990s the author co-sponsored a comparative survey of these structures
along with all respective owners. The resulting report by Columbia University
summarized cable conditions, design, construction and maintenance practices. Key
findings were presented by Betti and Yanev (1999).
The National Cooperative Highway Research Program (NCHRP) at theTransportation Research Board (TRB) expanded the study to all parallel wire cablesupported bridges nationwide and published NCHRP Report 534 by Mayrbaurl andCamo (2004) of Weidlinger Assoc. The report addressed twenty nine suspension
bridges, constructed in North America by the aerial spinning method up to the year
2000 and two (Newport, R.I. and William Preston Lane Jr., MA.) built with shopfabricated parallel wire strands. The twenty one (shorter) spans supported by helicalstrand cables were not considered in this study. NCHRP Report 534 recommended
further investigation of the behavior of cables under controlled and actual field
conditions.
Wires: condition, deterioration and failure
The replacement of suspension chains and eye-bars with parallel wires orhelical strands radically improved the internal redundancy of suspension cables. The
wires in the investigated cables (Williamsburg excepted) are galvanized, have anapproximately 5 mm diameter and their strength is around 1515 mPa. In a load-freestate, they assume a curvature with roughly 1 m diameter, which indicates their stateof bending under working conditions. The non-galvanized wires of the Williamsburg
Bridge failed when corrosion reduced their cross-section (Fig. 2 - A and - B). Thismode of wire failure is atypical, as well as relatively simple and has not attracted
much interest. Failures of non-galvanized helical strands have been reported in detailby Virlogeux (1999) and Kretz et al. (2006), along with the ensuing cablereplacements. In contrast, high-strength galvanized wires have proven susceptible to"flat and invert" breaks, and the rarer spiraling breaks (Mayrbaurl and Camo, 2004).
Flat breaks are believed to develop as follows:- The approximately 0.05 mm zinc coating oxidizes and fails over a small areas of thewire surface.
- The exposed steel begins to oxidize, causing surface irregularities, such as pitting
and, consequently, stress concentration.

253

- Cracks develop transversely to the wire surface, further concentrating the stress.
- The cracks propagate at an angle towards the center of the wire until the area iscritically reduced and the wire breaks normally to the axis. The sequence raises thefollowing critical questions:- Quantifying and qualifying wire corrosion.
The strength evaluation of a cable is based on visual inspections which, in turn
must estimate the state of corrosion. To facilitate these estimates, 4 stages of wire
corrosion were described by Hopwood and Havens (1984) as follows:Stage 1 - Spots of zinc oxidation on the wires;Stage 2 - Zinc oxidation on the entire wire surface;
Stage 3 - Spots of brown rust covering up to 30% of the surface of a 3 to 6 inch
(75 to 150 mm) length of wire;Stage 4 - Brown rust covering more than 30% of the surface of the 3 to 6 inch lengthof wire.

Figure 2. Suspension cable wires, z-shaped wrapping wire, and 127 wire strand
Despite the lack of phenomenological backing, this rating system endures, as
do all visual inspections. NCHRP Report 534 proposed a model linking the visual
findings, classified in the 4 - stage system to an estimate of the number of cracked andbroken wires, and ultimately, the cable strength.
- Does the oxidized zinc contribute the embrittlement of the steel?
Mayrbaurl and Camo (2004) point out that much depends on the further
reactions of the resulting zinc oxide (ZnO). Those, in turn depend on the environment
and, particularly on the type of humidity. Zinc carbonate (ZnCO3) can form aneffective protective film, whereas zinc hydroxide (Zn(OH)2) easily dissolves, leading
to the formation of carbonic acid (H2CO3), which becomes a source of embrittlinghydrogen.

254

- Is there a threshold level of zinc depletion and steel corrosion beyond which cracks
begin to occur spontaneously?
To some extent that would depend on the level of stress in the "as-built" cable.
Modeling that stress however, besides relying on the uncertain state of the unstressedwires, also depends on their stress level, subject to diverse uncertainties.
Wire stresses
Suspension cable wires are not in uniaxial tension as is generally assumed in
calculating safety factors. High-strength wires are manufactured by the cold-drawn
method by extruding them through progressively smaller openings and thus modifying
the molecular shape of the original mild steel. Eventually the wires of the desired
diameter (roughly 5.1 mm) are dipped in molten zinc for protection against corrosion.
The galvanized wires cool off in a permanently curved shape with a diameter ofapproximately 2 m. Consequently, the wires experience bending stresses in order toconform to the shape of the cable. Mayrbaurl and Camo (2004) estimate (based on Xray diffraction tests) that the straightening produces bending stresses of up to + 240
mPa (36 ksi). Eliminating a curvature with radius R induces in a wire with radius rwire
bending moment M as follows:
M=EI/R(1)

The corresponding maximum bending stress σ is:
σ = M / S = E rwire / R(2)

where: I = π rwire 4 / 4 is the section moment of inertia.
S = I / rwire is the section modulus.
For rwire = 5 mm and R = 2000 mm in Eq. (2), σ = 500 mPa (75 ksi), matching
the uniform working stress level for which many suspension cables are designed. Even
after many years of service, suspension bridge wires extracted from a cable regain someof their original curvature, testifying to inelastic deformation. Wires invariably crack on
the side where straightening has produced tension. Bending is extreme in the traditionalanchorages which use sheaves of relatively small diameters. This has been a primaryreason for the recent shift, lead by Japan, to prefabricated straight wire strands,
originally developed under a U.S. patent (Fig. 2 – F).
Residual stresses
Residual stresses would be caused by the extrusion process and by the hotdipping into zinc. The steel surface under the zinc is irregular. Therefore stressconcentration and even cracks may exist in the wires before the zinc coating fails.
From wires to a cable: condition, deterioration, maintenance and repair
The knowledge gained from individual wires must be fully used in evaluating
the condition of suspension cables. Whereas wire failures have provided
understanding of the limits of the material, a cable failure model does not exist andallowing for one to develop spontaneously is not acceptable. Consequently,

255physical and statistical, deterministic and probabilistic, analytic and empiricalmethods are comtinually developed and combined. All stages of the structural life areclosely examined.
Design and construction
Over the 20th Century parallel wire suspension cables (Fig. 3) evolved as follows:

(a)
(b)
(c)
Figure 3. Cables: (a) wedging; (b) cable band with red lead; (c) Akashi – Kaikyo
The number of suspension cables evolved from 2 to 4 and back to 2. Examplesof the 4 cable configuration include the East River bridges, the George Washington
and the Verrazano in New York City. Notable among the 2 cable bridges are theAmbassador, Macinac (1,158 m), Bay Bridge and Golden Gate (1,280 m). Four cables
were originally contemplated at the Akashi-Kaikyo, however the 2 cable cross-section
was selected because of its superior aerodynamic properties. The proposed 3,000 m
span at the Messina Straights assumes 4 cables. The global redundancy of the 4cable configuration may have saved some bridges from demolition. For his bridge at
Oporto, D. Steinman anticipated a second pair of cables carrying added traffic. That
modification was completed in the 1990s. At Pont de Tancarville the two original
cables were replaced by two pairs, allowing for a future replacement of each pair by asingle cable at the original location (Virlogeux, 1999). The two cables of theStorabaelt Bridge in Denmark (1996, main span 1,624 m) were built by the traditional
air-spinning method with 18,648 wires of diameter 5.37 mm and minimum strength of
1,570 N/mm2 (Fig. 2 - C). The Akashi – Kaikyo (1998, main span 1991 m) cables arebuilt with shop-fabricated parallel wire strands, composed of 127 straight galvanizedwires with yield up to 1800 N/mm2 (Figs. 2 - D and F, Fig. 3 - c). Each of the two
cables have 290 hexagonal strands (Figs. 2 – F and 3 - c).
Suspenders
Once spaced at roughly 7 m, suspenders are now spaced at about 20 m. The
increased distance between suspenders affects primarily their own behavior as well asthat of the bridge superstructure. The cables, however are also influenced in at least
two ways. The cable bands which improve the behavior of broken wires and the256compaction of the cable are reduced. Consequently they transmit greater concentrated
loads to the wires. As the distance between the cable bands increases so does thebending moment introduced by them into the cable.
The traditional corrosion protection of suspenders has been galvanization andpaint. Stays are protected in various encasements, primarily cement grout. The
limitation of the latter method, however have prompted Japanese designers to use
parallel wire suspenders and stays with rubber vulcanization, as at the Akashi Kaikyo and Tatara Bridges.
Wind loads
The potential for aerodynamic instability demonstrated by Tacoma and othersuspension bridges led to extensive testing of models of all major suspension andcable-stayed bridges in wind tunnels. For the Akashi – Kaikyo Bridge, a 1/100 scalemodel was tested in a wind tunnel built expressly for that purpose.
The inclined suspenders used on many bridges in Europe, including the recordbreaking Humber in Wales (1981, 1,410 m main span) proved even more vulnerable
to wind loads and fatigue. After relatively short useful lives, such suspenders were
replaced at the Severn and Brotonne bridges while their sockets were modified. Wateron the surface of stays was found to modify their aerodynamic response, causing theso-called galloping oscillations. The phenomenon is mitigated by surface
obstructions to running water. Dampers are added on longer suspenders and stays.
Clamping length
The overall stress in cables is estimated by cutting wires and measuring theirretraction. That retraction, however is constrained by the friction with adjacent wires
and is therefore limited to the so-called clamping length of the wires. The clamping
length is very important in estimating the contribution of broken wires at longer
distances from the fracture points. It is often assumed that tension in a broken wire isrestored over the distance of 3 cable bands. Apart from the hypothetical nature of thatassumption, it clearly depends on the distance between cable bands and on the level ofcompaction of the cable. Thus the local condition of the wires must be assessed
jointly with the global condition of their totality as a cable.
Safety factor
This term is somewhat discredited, because of the uncertain assessments of bothstress and resistance levels discussed in the preceding section. Nonetheless, the ratio
between the ultimate load of the cables at yield and the maximum expected load
during its service has declined from over 4 to 2.2. It is argued that 90% of all loads onlong-span bridges consist of the relatively constant dead load. Furthermore, theeffective stiffness of the cables increases proportionally to the cube of the stressdivided by the square of the span length. The formula, attributed to Tischinger and to
Ernst, can be written as in Eq. 3.
1/Eef = 1/E + (γL)2 / (12 σ3)

257

(3)

where: Eef - effective modulus of elasticity of the cable;
E - modulus of elasticity of the steel;L - span length;
γ - specific weight of the cable;
σ - stress in the cable.
Thus longer spans achieve the desired stiffness at the expense of higher
stresses in the wires, implying an increased likelihood for "stress-corrosion" andhence a lower tolerance for deterioration.
Cable protection
Cable protection once consisted of linseed oil introduced into the cable voids, a
red lead paste coating over the wires, plastic wrapping over the lead, spiral wire
wrapping on top and paint over the spiral wire (Fig. 3 – a and b). Because ofenvironmental objections, lead is being substituted by zinc paste. The long-term
effects of this modification are yet to be observed, because lead is passive whereas
zinc oxidizes. The benefits from the spiral wire wrapping have been disputed and it
has been replaced by polyethylene wrapping at some bridges.
On several record-breaking bridges in Japan, such as the Kurushima and
Akashi-Kaikyo, managed by the Honshu-Shikoku Bridge Authority, pressurized dry
air is injected under the cable wrapping. The system is now under consideration as arehabilitation feature on older bridges. Z-shaped wires have been used for wrapping,as in Fig. 2 - E.
Compaction
Compaction is a form of cable protection, because it impedes the penetrationof water and increases the clamping effect. Air-spinning achieves a relatively unevencompaction of 75%-80%. In a perfectly compact cable all wires except the external
ones are in contact with 6 adjacent wires, forming a hexagon. For T perfectlycompacted concentric layers, the net wire to gross cable ratio of areas can becomputed as in Eq. 4.
2π r2 [3T(T + 1) + 1] / r2 [(2T + 1)2 33/2] ≈ 1.211 x 3/4 ≈ 0.907(4)

In cables built by the traditional air-spinning method 80% compaction isrealistic, but variations are quite broad.
Anchorages
The anchorages of air-spun cables (Fig. 4 - a) feature three critical transitions:- From a compacted cable to splaying strands. In this area wires no longer havethe benefit if clamping effects and are fully exposed to humidity. As a result entire
strands have been lost to corrosion in anchorages, and re-anchorings have beennecessary (Fig. 4 - a).
- From strands to eye-bars. The bending over the pins of the eye-bars can cause
yield in the wires, but few breaks have been noted in these areas.

258

- From exposed eye-bars to concrete encasement. Eye-bars corroded
significantly at that juncture and have been replaced by new anchoring systems, forexample at Manhattan Bridge. Most anchorages (including old ones) are being
equipped with de-humidification systems.

(a)(b)
Figure 4. Anchorages (a) air-spun cable with re-anchoring; (b) prefabricated cable
Uncertainty
For engineering purposes uncertainty has been separated by ISO (1995) into
randomness (for example of natural phenomena), vagueness (as in condition
evaluation) and ignorance (of actual conditions). Ang and De Leon (2005) recognize
aleatory and epistemic uncertainties, associated with random natural variables anddeterministic risk-informed decisions, respectively. In either classification, all types of
uncertainties are present in bridge cables. Report NCHRP 534 treats probabilisticallythe key parameters of the cable model as follows:
Wire condition
Acoustic emission has been used to detect wire breaks of cables in use. The
labor-intensive and intrusive unwrapping and wedging (Fig. 3 - a) remains the only
reliable source of information about the condition of the wires. NCHRP 534 proposes
methods of projecting the number of wires in each of the four states, based on thelimited findings of such inspections. The extrapolations depend on the sizes of thecables and the sample, and the observed conditions.
Cable strength
NCHRP 534 proposes three models as follows:
Ductile wire. Strain increases incrementally. A wire reaching yield carries thecorresponding stress until the rest of the wires reach that point and they all failsimultaneously. The cable strength is the average strength of the wires multiplied bytheir number. This model has a limited application to cables (possibly such as those of
the Williamsburg) where loss of section, rather than cracking is critical.

259

Brittle wire. Each wire fails at its tensile strength limit. A step-wise stressstrain diagram results, as the number of wires declines. The model tends tounderestimate the cable strength by as much as 20%.
Limited ductility model. This more elaborate model recognizes that when thefirst wire breaks, the force in the remaining wires does not change because the change
in the overall strain is negligible. As the ratio of broken to active wires grows, so does
the strain rate. The process is dynamic.
Health monitoring
Wind tunnel testing has become standard practice on all suspension and cablestay bridges, however scale models cannot fully simulate the actual structural
response. Consequently, dynamic analysis must follow a redundant path, seeking
acceptable convergence of analytic and experimental results, as has been the casesince the origin of bridge building. The process does not stop with the completion of
construction. Monitoring systems collect “real-time” data including dynamic
response, stresses, temperatures, traffic weigh in motion, wind velocity.

Figure 5. Cable model at Columbia University
New suspension bridges are heavily instrumented with various devices,
measuring acceleration, vehicular weight, wind speed, temperature and otherenvironmental and structural parameters. Global positioning systems (GPS) canmonitor the movements of large structures with a satisfactory accuracy. Recently
Manhattan Bridge was scanned with interferometric radar by IDS Ingeneria of Pisa,
Italy and with GPS by Columbia University with closely agreeing results. The time260

has come to introduce state of the art instrumentation into the cables and to employ it
for their continuous health monitoring. Under a current project sponsored by FHWA,
Columbia University researchers are investigating the methods for non-invasive
monitoring of cable condition. The research encompasses the testing of a cable model(Fig. 5) under controlled conditions and transferring the monitoring technology to asuspension bridge for field verification. The cable model is instrumented with 76
sensors of different types, placed along three diameters, inclined at 60o angle withrespect to each other, so to have a dataset that represents the central conditions for the
6 sectors in which the cross-section is typically subdivided. These sensors measuretemperature, relative humidity, pH and corrosion rate. The sensor allocation is shownin Fig. 6. The 16 temperature/relative humidity sensors were evenly spaced along thethree diameters in order to have measurements for the temperature and humidity
variations inside the entire cross-section. Corrosion rate sensors used different
principles to measure corrosion rates: some use Linear Polarization Resistance, someothers are Bi-metallic and some others, like the Couple Multiple Array Sensors,
measure the number of electrons that flow from anodic to cathodic zones.

Figure 6. Sensor allocation in the cable model

261

The objective is to identify methods suitable for monitoring the condition ofcables without the need to unwrap and wedge them. The instruments selected during
the laboratory testing phase will be installed in one of the cables of Manhattan Bridgefor field testing in 2010. Thus the design, construction and service of suspension
bridges reaches for new levels of theory and application. Once again theory mustcatch up with practice.
Acknowledgment
The views presented in this article are those of the author and do not express
the position of any organization or agency.
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Thoft-Christensen, P. and M. J. Baker (1982) Structural Reliability Theory and Its
Applications, Springer - Verlag, Berlin.
Virlogeux, M. (1999) Replacement of the Suspension System of the Tancarville
Bridge, Transportation Research Record 1654, National Academy Press, Washington,
D.C., pp. 113 - 120.
Von Karman, T. and M. A. Biot (1940) Mathematical Methods in Engineering,
McGraw-Hill, New York.
Yanev, B. (2007) Bridge Management, John Wiley & Sons, Inc, Hoboken, N.J.

262

25th US-Japan
Bridge Engineering Workshop

Session6

Seismic Performance Evaluation
Dynamic Response Analysis of Bridge Under Seismic Loading Including Collision
By Eiki Yamaguchi, Atsumi Ryuen, Keita Yamada, and Ryo Okamoto
Effects of Near-Fault Vertical Accelerations on Highway Bridge Columns
By Sashi Kunnath and Huiling Zhao
Seismic Design of Multi Span Continuous RIGID-FRAME Bridge with Prestressed
Concrete Box Girder
By Chiaki Nagao, Yasushi Kamihigashi, Akio Kasuga, and Kenichi Nakatsumi

Page Left Blank

DYNAMIC RESPONSE ANALYSIS OF BRIDGE UNDERSEISMIC LOADING INCLUDING COLLISION
Eiki Yamaguchi1, Atsumi Ryuen2, Keita Yamada2, Ryo Okamoto2
Abstract
In the Mid Niigata Prefecture Earthquake in 2004, the damage of a bridge pierwas found mitigated by the collision between a girder and an abutment. Therefore, inthe seismic response analysis, it is important to include the effect of the collisions. A
simple and practical method of simulating the collision in the analysis is theintroduction of a spring where the collision occurs. However, the constant of such acollision spring is yet to be formulated well. In this study, appropriate collision-spring
constants are investigated for simulating the collisions between girders and between agirder and an abutment. The response analysis of a bridge under seismic loading is thenconducted to see the influence of the collision.
Introduction
The collision between a bridge girder and an abutment restricts the movementof the girder, which in turn reduces the deformation of a bridge pier, often considerably.
This implies that the collision can help mitigate seismic damage of the bridge. In fact,
in the aftermath of the Mid Niigata Prefecture Earthquake in 2004 we studied adamaged bridge pier (Photo 1) and concluded that the damage could have been muchworse if not for the collision between a bridge girder and an abutment (Kosa et al.
2005). Thus the collision plays an important role in the behavior of a bridge during
earthquake, and without including the effect of the collision, the analysis of a bridge
behavior during earthquake can be quite misleading.
A simple and practical method of simulating collision in the analysis is theintroduction of a spring between two bodies that collide with each other. This spring iscalled a collision spring and activates only when the two bodies collide with eachother; the constant of the collision spring is kept equal to zero when the two bodies are
not in contact. The effectiveness of this approach hinges on the behavior of thecollision spring. In short, the value of the collision-spring constant is important.
Nevertheless, the appropriate value has not been clarified except when the two bodies
are identical (Kawashima 1981).
In the present study, the appropriate value of the collision-spring constant
between two bodies having different stiffnesses is investigated. The response analysis
of a bridge under seismic loading is then conducted to demonstrate the influence of thecollision.

1
2

Professor, Dept. of Civil Engineering, Kyushu Institute of Technology, Kitakyushu
Student, Dept. of Civil Engineering, Kyushu Institute of Technology, Kitakyushu

265

(a) Overview(b) View around girder end

Photo 1. Bridge damaged during the Mid Niigata Prefecture Earthquake in 2004.

GIRDER 1

COLLISIONSPRING K

k* = 1 (kN/m)
GIRDER 2
E2A2

E1A1
L2

L1

Figure 1. Model A.
COLLISIONSPRING K

k* = 1 (kN/m)
GIRDEREA

ABUTMENTSPRING ka

L

Figure 2. Model B.
Collision-Spring ConstantAnalysis Models
The analysis models, Models A and B, are shown in Figures 1 and 2. The twomodels are to study the collisions between two girders and between an abutment and agirder, respectively. The abutment is modeled by a spring and the girder is by beam
elements. The spring with a very small spring constant at the right end of the girder isintroduced to prevent the free movement of the girder.
The collision spring is placed between the bodies that would collide. The
behavior of the collision spring is shown schematically in Figure 3. The relative
displacement is the distance between the two bodies. The origin in Figure 3 represents

266

2000

FORCE

-u0

1500

RELATIVE DISPL.

1000

500

0

Figure 3. Collision Spring.

0

0.02

0.04 0.06 0.08
TIME (sec)

0.1

Figure 4. Acceleration.

the initial state. The figure implies that when the relative distance is smaller than -u0,
the collision spring is activated: its spring constant becomes non-zero, simulating thephysical contact of the two bodies. The value u0 corresponds to the initial gap between
the two bodies.
Going through the dimensions of some existing composite girder bridges inJapan, the following values are employed for the present study:
Model A
E1A1/L1 = 5.0×105 kN/m
E2A2/L2 = 5.0×106 kN/m
Model B
EA/L = 5.0×105, 1.0×106, 5.0×106 kN/m
ka = 1.0×105, 3.0×105, 1.0×106, 3.0×106, 1.0×107, 3.0×107, 1.0×108, 3.0×108 kN/m
where E is Young’s modulus, A the cross-sectional area, L the girder length and ka is
the stiffness of the abutment.
The initial gaps u0 between the two girders and between the girder and theabutment are all set equal to 0.15 m. The acceleration shown in Figure 4 is applied in
the longitudinal direction for the sake of simplicity. The damping coefficient of thegirder is 0.02. Material nonlinearity is not considered. The increment of time for time
integration in the present dynamic analysis is Δt =1/50000 sec and 100 beam elements
are used for modeling each girder.
Numerical Results and Discussion
Model AThe collision spring has been employed to carry out the dynamic response
analysis of a bridge that involves collision. Yet the appropriate spring constant, the267slope in the negative region beyond -u0, has not been defined well even though it islikely to influence numerical result. To the best of the authors’ knowledge,
Kawashima’s work (1981) is the only one that deals with this issue squarely. He
studied the collision of two identical bars and came up with the following proposal:
k= nEA/L(1)

where n is the number of elements. To be noted, the usage of the finite element method
is presumed and Equation (1) indicates the collision-spring constant may be set equal
to the longitudinal stiffness of the element.
This equation can be used right away if the two bodies yield the same k. But
oftentimes that is not the case. The present analysis is one of those cases: two different
constants of the collision spring are obtained by Equation (1). Therefore, assuming thetwo collision springs connected in series, the following formula for the collision-spring
constant K is proposed:
K=

1
1
1
+
k1 k 2=

k1k 2
k1 + k 2(2)

where k1 and k2 are the values obtained by Equation (1) for Girders 1 and 2,
respectively.
To investigate the validity of Equation (2), the dynamic analysis of Model A is
conducted with various collision-spring constants K in the range of 1.0×106 - 5.0×107
kN/m.
The results are presented in Figure 5. A small collision-spring constant leads tolarge relative displacement in the negative region: large overlapping of the two bodies
occurs, which is a fictitious phenomenon arising from the approximation of thecollision behavior in the numerical simulation. The overlapping can be suppressed byusing a larger collision-spring constant. But the large collision-spring constant results
in a peculiar phenomenon, vibration in the relative speed. Moderate magnitude of thecollision-spring constant, not too small and not too large, needs be employed. Judging
from the results in Figure 5, it may be realized that the collision-spring constant K in
the range of 5.0×106 - 5.0×107 kN/m is acceptable. Since Equation (2) gives 9.0×106,
this formula can be considered valid for the collision-spring constant between twogirders having different stiffnesses.
Model B
In general, the stiffnesses of the girder and the abutment are not the same andthe dynamic behavior of the abutment during earthquake is very different from that of
the girder. No good guidelines are available for the collision-spring constant K between
the girder and the abutment.

268

4

6

K=1.0x10

K=1.0x1067

K=5.0x102

10K=5.0x10

K=9.0x10

K=1.0x10

K=5.0x106

K=5.0x107

K=9.0x106

5

6

7

6

K=1.0x10

15

7

0
0

-5
-10

-2

-15
-4

0.02

0.03
0.04
TIME (sec)

0.05

-20
0.02

0.06

(a) Relative displacement.

0.021 0.022 0.023 0.024 0.025
TIME (sec)(b) Relative speed.

Figure 5. Results of Model A.
Since the dynamic movement of the girder is much bigger, the collision-spring
constant may be determined by Equation (1) solely with the girder stiffness The
influence of the abutment on the collision-spring constant is ignored. The
collision-spring constant thus obtained is denoted by k. Referring to the result of Model
A, Equation (2) may be used with k1 computed by Equation (1) with the girder stiffness
and k2 = ka. This collision-spring constant is K0.
With various values of collision-spring constants, the dynamic response
analysis of Model B is conducted. The range of the acceptable collision-spring
constants are found through this analysis and shown in Figure 6 as line segments. In
this figure, k and K0 mentioned in the previous paragraph are also plotted and both arefound unacceptable.
It is also noticed in the numerical results that the maximum collision-spring
constant Kmax in each acceptable range varies almost linearly with ka in thisdouble-logarithmic graph of Figure 6. Figure 7 further presents the maximum
acceptable collision-spring constants for all the cases, showing that the lines
connecting the maximum acceptable collision-spring constants associated with thesame values of k1 are parallel to each other.
Based on these observations, the following formula is constructed:
log Kmax = log ka +α(3)where α is dependent on k1 and the regression analysis yields

α = −0.945 × log k1 + 11.2

269

(4)

12

10

1011

1011

1010

Kmax (kN/m)

COLLISION-SPRINGCONSTANT (kN/m)

1012

109
108
107

k

106
105
4
10

5

6

7

10
10
k (kN/m)

8

10

10

109
5

8

EA/L=5.0x10 (kN/m)

7

EA/L=1.0x10 (kN/m)

10

6

10

K0
10

10

10

EA/L=5.0x106 (kN/m)

6

10

9

4

10

a

Figure 6. Results of Model B.

5

10

6

7

10
10
ka (kN/m)

8

10

9

10

Figure 7. Maximum acceptable collisionSpring constants.

Then the following formula may be proposed to estimate the collision-spring constantK:
log K = log ka +0.9α(5)Response Analysis of Girder Bridge
Referring to an existing bridge, a bridge model is constructed and analyzed.
Figure 8 shows the schematic of the bridge: the bridge is a 7-span cantilever girder andthe length is 189.45 m. In the figure, M, F, m and f stand for a movable bearing support,
a fixed bearing support, a movable hinge and a fixed hinge, respectively. An abutment
is assumed at each end of the bridge.
The superstructure is a composite girder with four steel I-shaped girders and aconcrete deck. The web plate is constant while the upper-flange width varies from 290
to 490 mm and the upper-flange thickness is from 12 mm to 25 mm; the lower-flange
width varies from 200 to 540 mm and the upper-flange thickness from 12 mm to 25 mm.
The bridge piers are of a rectangular cross section with 2.2 m × 10.6 m at the top and3.6 m × 13.6 m at the bottom. The nonlinear material behaviors of steel and concrete
shown in Figure 9 are assumed with the yield stress of steel σy and the compressive
strength of concrete σck equal to 360N/mm2 and 21N/mm2, respectively. The behaviorof the foundation is modeled by springs whose constants are determined by Design
Specifications of Highway Bridges (Japan Road Association 2002). The
collision-spring constants are determined by the formulas given above.
The seismic wave in the form of acceleration recorded during the Mid Niigata
Prefecture Earthquake in 2004 is used in the analysis (Figure 10). The initial gaps are
assumed 0.25 m. Two analyses are conducted: one ignores the collision effect and theother includes it.
Figures 11 and 12 summarize the results of the movements of the girders and270

189450

F

z
y

x

P1

P2

P4

P3

M

13140

F

14930

M

13040

F

12530

M

12920

A1

13030

24700 6000 16000 6000 16000 6000 16000 6000 16560 6000 15490 6000 24700
m
m f
m f
f

P5

F

M

A2

P6

(a) Side view.
9800
3@2600=7800

1000

y

G2

G4

1800 220

1000

z
x

G1

G3

(b) Cross section.
Figure 8. Bridge model (Units: mm).

(a) Steel.

(b) Concrete.

Figure 9. Material behavior.
piers. Figure 11 shows the maximum horizontal displacement at four points in the
superstructure: the closest point to A1, the mid-span, the closest point to A2 and thepoint right above P3. Figure 12 presents the maximum horizontal displacements of P1,
P3, P4 and P6. It is clearly observed from these figures that the collisions suppress the
movement of the bridge, possibly reducing the damages of the piers.

271

2000
1500
1000
500
0
-500
-1000
-1500

0

5

10
15
20
TIME (sec)25

30

Figure 10. Acceleration recorded during the Mid Niigata Prefecture Earthquake in
2004.
INITIAL GAP 0.25m
NO COLLISIONIN ITIAL GAP 0.25m
NO CO LLISION

0.55

0.55

0.5

0.5

0.45

0.45

0.4

0.4

0.35

0.35

0.3

0.3

0.25

0.25

0.2
0.15

0.2
A1 MID-SPAN A2

P3

0.15

LOCATION

P1

P3

P4

P6

Figure 11. Maximum horizontal
Figure 12. Maximum horizontal
displacement of superstructure.
displacement of piers.
Concluding Remarks
To include the effect of the collisions, the so-called collision springs are often
used. Yet the appropriate constant of the collision spring is not available in theliterature except when the two bodies are identical. The present study tackled this issue
and, through numerical study, it proposed the formulas for acceptable collision-spring
constants between girders and between a girder and an abutment. The dynamic
response analysis of a girder bridge was then conducted and the significant influenceof the collisions was indeed observed. Therefore, it is crucial to simulate collisions
when the dynamic response analysis of a bridge is conducted. To that end, the formulas
for acceptable collision-spring constants should be useful.
References272

Kawashima, K.: An analytical model of contact and impact in dynamic response
analysis, Technical Note, Journal of JSCE, Vol. 308, pp. 123-126, 1981.
Kosa, K., Yamaguchi, E., Inokuma, Y., Tasaki, K.: Investigation and analysis ofbridges and other structures damaged during the Mid Niigata Prefecture Earthquake in
2004, Proceedings of the 8th Symposium on Ductility Design Method for Bridges, pp.
35-40, 2005.
Japan Road Association: Design Specifications of Highway Bridges: Part V Seismic
Design, Maruzen , Japan, 2002.

273

blankEFFECTS OF NEAR-FAULT VERTICAL ACCELERATIONS ON HIGHWAY
BRIDGE COLUMNS
Sashi K. Kunnath1 and Huiling Zhao2
Abstract
The effects of vertical ground motions on the seismic response of ordinary
highway bridges are investigated. Nonlinear simulation models with varying
configurations of an existing bridge in California are developed for use in a detailed
parametric study. The models are subjected to earthquake motions with and without
vertical accelerations. Results indicate that vertical effects lead to significant variations
in axial force demand in columns which can result in: fluctuations in moment demands
at the face of the bent cap, amplification of moment demands at the girder mid-span,and changes to moment and shear capacity of the column. In the second phase, theeffect of vertical motions on shear demand and capacity of bridge columns are
examined.
Introduction
The Seismic Design Criteria (SDC) used by the California Department ofTransportation for ordinary standard bridges states that vertical effects should beconsidered on sites where the peak rock acceleration is expected to be more than 0.6g.
However, the procedure to evaluate vertical effects is rather simplistic: a separate
equivalent static vertical load analysis should be carried out under a uniformly
distributed vertical load of 25% of the dead load applied in the upward and downward
directions, respectively (CalTrans 2006).
Recent earthquakes have revealed that the ratio of vertical to horizontal peakground acceleration can be larger in near-fault records than far-fault records. Hence it
has become necessary to reexamine the consequences of vertical motions on typicalhighway bridges. The characteristics of vertical motions and the effect of verticalaccelerations on bridge structures have been investigated by several researchers(Saadeghvaziri and Foutch 1991; Bozorgnia and Niazi 1993; Broekhuizen 1996;
Papazoglou and Elnashai 1996; Yu et al. 1997; Gloyd 1997; Collier and Elnashai 2001;
Button et al. 2002). Among other findings, these studies conclude that the variation of
axial forces due to vertical excitations can influence both the moment and shear
capacity of the section and also increase tensile stresses in the deck. An evaluation of
the characteristics of response spectra of free-field vertical motions recorded during the1994 Northridge earthquake by Bozorgnia et al. (2004) found the vertical to horizontal(V/H) response spectral ratios to be strongly dependent on period and site-to-source
distance.
1

Professor, Department of Civil and Environmental Engineering, University ofCalifornia, Davis, CA
2
Visiting Graduate Student, University of California at Davis & PhD Candidate,
Tongji University, China

275

Important Characteristics of Vertical Accelerations
A preliminary study of 65 near-fault earthquake records with horizontal PGA
greater than 0.5g indicates that:• The predominant period of the vertical ground motions are smaller than thecorresponding horizontal component (Figure 1).
• The ratio of vertical-to-horizontal PGA decreases gradually with increasing
fault distance, therefore, the vertical component of ground motions will be more
severe for near fault ground motions.
3.0

VERTICALHORIZONTAL

2.5

SA/PGA

2.0
1.5
1.0
0.5
0.0
0.0

0.5

1.0
1.5
PERIOD (SEC)

2.0

Figure 1: Characteristics of near-fault horizontal and vertical motions
Summary of Phase I Study
An existing two-span overcrossing was selected to represent a typical ordinary
highway bridge in California. The computer model used in the simulations is shown in
Figure 2. The superstructure consists of a reinforced concrete box girder supported by
two circular columns with a diameter of 1.78 m. The width of the girder was unchanged
at S = 8.8 m and the column height was fixed at H = 8.5 m.

L
L 2
Support
y
S
S

x

L1
L
z
H
H

Figure 2 – Simulation model of typical two-span bridge

276

The end conditions both at the abutments and at the bottom of the columns weremodeled using spring elements whose properties were determined using SDC(CalTrans 2006) guidelines. The superstructure was modeled both as elastic elementsin the initial phase of the study and later as inelastic elements to examine the effects ofinelasticity on reinforcement yielding. Potential inelastic regions of the columns weremodeled using fiber hinge elements with prescribed plastic hinge lengths. Axial force– moment interaction was therefore included in the simulations. Additional details of
the bridge and the simulation models are reported in Kunnath et al. (2007). In order toinvestigate the effect of vertical accelerations on a wider range of vertical frequencies,different bridge configurations were created by modifying the span lengths, L1 and L2.
Table 1 presents the fundamental dynamic properties of the selected configurations.
Table 1 – Properties and periods of bridge configurations
Simulation
Model #TV (s)TL (s)TT (s)

1
2
3
4
5
6

0.19
0.12
0.30
0.37
0.45
0.24

0.32
0.27
0.43
0.53
0.62
0.35

0.55
0.46
0.64
0.68
0.75
0.59

2.00

Mean Horizontal

Spectral Accel. (g)1.75

Mean Vertical

1.50
1.25
1.00
0.75
0.50
0.25
0.00
0.0

0.5

1.0

1.5

2.0

2.5

Period (sec)Figure 3 –Spectra of the horizontal component of ground motions scaled to match the
ARS curve (Magnitude 8.0, PGA 0.5 g and site class D) at fundamental longitudinal
period of base configuration, and corresponding vertical spectra
Ground Motions
Following a preliminary set of analyses, a reduced subset of 29 near-fault
records that produced the largest demands on the bridge was selected for detailed
evaluation. All ground motions were scaled to match the ARS spectrum (CalTrans2006) at the longitudinal period of each bridge configuration. Figure 3 displays typical277

spectra of the horizontal component of the ground motions scaled to match the ARS
curve for ground motions with a PGA 0.5g and site class C together with thecorresponding vertical spectra.
Effect of Vertical Acceleration on Column Axial Force
Figure 4 summarizes the variation of the normalized axial load as a function of
the vertical fundamental period. Both the maximum and minimum axial force
experienced by the column in each simulation is recorded. The amplification of theaxial load in the column is not a source for concern since the nominal axial load
capacity of the columns is adequate to resist these forces without damage. However, thevariation in the axial force on the column may result in significant changes in the
moment and shear capacity of the column.

Axial Force (H+V)/DL

6.00
Mean (max & min)

4.00
2.00
0.00
-2.00
-4.00
0.10

0.15

0.20

0.25

0.30

0.35

0.40

0.45

Structural Vertical Period (s)Figure 4 – Variation of column axial force demands with vertical period for unscaled
ground motions
Effect of Vertical Acceleration on Span Moment
Figure 5 shows the variation of the normalized moment demand (ratio of
moment demand to the moment demand due to dead load only) at mid-span of the left
girder as a function of the fundamental vertical period. The results highlight thesignificant effects of vertical motions on the moment demands in the longitudinal
girders. It should be pointed out that the girders were modeled as elastic elements inthese simulations. Since the negative moments far exceed the available capacity, thesimulations were repeated using inelastic elements for the girder. Peak strains were
found to vary up to 12 times the yield strain.

278

Span Moment (H+V)/DL

Capacity
Mean (H+V)

8.0
6.0
4.0
2.0
0.0
-2.0
-4.0
0.10

0.15

0.20

0.25

0.30

0.35

0.40

0.45

Structural Vertical Period (s)Figure 5 - Variation of moment demand at the mid-span with vertical period for
unscaled ground motions
Effect of Vertical Acceleration on Column Moment
The large increases in the axial compression in columns due to verticalaccelerations (Figure 4) can lead to significant increases in the column moment
capacity. Although this may suggest conservatism in the design, it may result in theshifting of the potential plastic hinge zone from the top of the column to the girder
which is an undesirable situation that is supposed to be avoided by the requirements of
SDC-2006.
Phase II Results
In reviewing some of the earlier work on the effects of vertical motions on
structures, some investigators have raised the issue of shear demand and capacity inbridge columns due to changes in the axial force demands. Since the bridge
configurations used in the Phase I study consisted of two-column bents and single
columns with very large shear span ratios, shear demands were generally not critical.
Hence a new study was initiated to identify critical bridge configurations that might beprone to shear damage due to vertical effects.
The typical configuration selected for this phase of the work is the Plumas
Bridge in California which is a three-span bridge with span lengths of 40.5m, 58.0m
and 40.5m. The heights of the as-built columns are approx 9 m each. In order to study
shear demands under strong near-fault motions, the column heights were varied togenerate a range of aspect ratios. A nonlinear simulation model, as displayed in Figure
6, was developed in OpenSees (2009). To ensure proper modeling of the torsional
properties of the deck, a three dimensional shell model of the bridge was created in
SAP-2000 and a series of elastic modal analyses were carried out on both systems tocalibrate the inertial properties of the superstructure of the line model.

279

Figure 6 – Simulation model of the Plumas Bridge used in Phase II Study

The model was subjected to a series of combined horizontal and verticalnear-fault motions. The earthquake records used in the simulations are listed in Table
2. Each set of records was scaled to match the SDC-ARS spectra (corresponding to amagnitude 8 event with PGA of 0.6g) at the fundamental transverse period of the bridge(which was estimated at 0.8 seconds). Scale factors were established for the larger of
the two horizontal components and this factor was used to scale the remaining two
components. A plot showing the mean spectra for all scaled records is summarized in
Figure 7.
Table 2 – Characteristics of selected near-fault records
Earthquake

Year

1. Gazli (USSR)
1976
2. Imperial Valley
1979
3. Morgan Hill
1984
4. Erzican (Turkey) 1992
5. Landers
1992
6. Northridge
1994
7. Kobe (Japan)
1995* Closest fault distance

StationKarakyr
Bonds Corner
Coyote Lake Dam
ErzincanLucerneRinaldi Rec Stn
KJMADistance*
(km)
5.46
2.68
0.30
4.38
2.19
6.50
0.96

280

PGA-HMAX
(g)
0.718
0.755
1.298
0.515
0.785
0.838
0.821

PGA-HMIN
(g)
0.608
0.588
0.711
0.496
0.721
0.472
0.599

PGA-Vert
(g)
1.264
0.425
0.388
0.248
0.818
0.852
0.343

3.0

SDC‐ARS
VerticalHorizontal‐Max
Horizontal‐Min

Acceleration/ g

2.5
2.0
1.5
1.0
0.5
0.0
0.0

0.5

1.0

1.5

2.0

2.5

Period/ s
Figure 7 – Spectra of ground motions used in Phase II study
For each simulation, the axial force, shear demand and shear capacity in thecolumn was monitored. Shear capacity at any instant in time was evaluated using
ACI-318 (2007) expressions:
Vn = Vc + Vs

(1)

In the above equation, Vc is shear strength provided by concrete, and Vs is shear
strength provided by shear reinforcement.
⎧ ⎛
Nu ⎞
′
when Nu > 0 (axial compression)⎪2 ⎜ 1 +⎟ f c bw d ,
⎜
⎟
2000
A
⎪⎪ ⎝
g ⎠
Vc = ⎨
⎪ ⎛
Nu ⎞
′
2
1
whenNu ≤ 0 (tension)+⎜⎪ ⎜ 500 A ⎟⎟ f c bw d ,g ⎠⎪⎩ ⎝Av f yt dVs =s
N u : factored axial force normal to the cross section

(2)

(3)

Figure 8 shows a typical response of the bridge column. The computed axial
force is normalized by the dead load which means that values below 1.0 indicate a stateof axial tension in the column. The shear capacity (which is a function of the axial
force) is superimposed on the demand plot so that the demand to capacity ratio (DCR)
can be ascertained. The particular case study presented in Figure 8 is for the Landers
record set which provided the maximum DCR among all ground motions considered.

281

Axial Force / Dead Load

6
4
2
0
‐2
‐4
0

5

10
TIME (SEC)

15

20

(a)

10000

SHEAR (kN)

5000
0
‐5000
‐10000

Shear Capacity‐15000

Shear Demand
0

5

10
TIME (SEC)

15

20Shear Demand/Capacity(b)
Figure 8 – Response of the bridge column with aspect ratio of 3.0 subjected to theLanders (1992) record: (a) Axial force variation; (b) Shear demand vs. available shear
capacity.

2.0

EQ.1

1.5

EQ.2
EQ.3

1.0

EQ.4

0.5

EQ.5
EQ.6

0.0
2.5

3.0

3.5

4.0

4.5

5.0

EQ.7

Column Aspect Ratio
Figure 9 – Effect of aspect ratio (column height/diameter) on shear DCR

282

Finally, a summary of the shear DCR for all simulations is presented in Figure
9. The record numbers correspond to the list identified in Table 2. It is evident thatshear damage is likely only for aspect ratios below 4.5. Given the fact that theexpressions used to estimate shear capacity are generally conservative, it is reasonable
to conclude that only columns with aspect ratios below 4.0 need further investigation.

Concluding Remarks
The main objective of the present study is to assess current provisions in
SDC-2006 for incorporating vertical effects of ground motions in seismic design ofordinary highway bridges. Results of the investigation suggest that highway
over-crossings with vertical periods close to the predominant period of the verticalcomponent of the motion are more vulnerable to vertical effects.
Findings also indicate that vertical ground motions significantly affect the axial
force demand in columns which in turn have an effect on moment demands at the
middle of the span. A separate study on the effects of vertical motions on shear
demands and shear capacity in the columns reveal that the aspect ratio (column height
to diameter) is a significant parameter that influences potential shear damage to thecolumn. It should also be noted that axial forces vary at much higher frequencies thanlateral forces. Hence the sudden shifts in shear capacity as the column goes fromcompression to tension may require further investigation. A shaking table test programon this issue is the subject of an ongoing investigation funded by Caltrans and being
carried out collaboratively between UC Berkeley and UC Davis.

Acknowledgements
Funding for this study provided by the California Department of Transportation(Caltrans) under Contracts No.59A0434 and No. 59A0688 is gratefully acknowledged.
The contributions of Li-Hong Sheng (Project Manager), Mark Yashinsky and several
Caltrans engineers throughout the project duration are also appreciated.

References
Bozorgnia, Y., M. Niazi. (1993). "Distance Scaling of Vertical and Horizontal
Response Spectra of the Loma Prieta Earthquake." Earthquake Engineering andStructural Dynamics, Vol 22, pp.695-707.
Bozorgnia, Y. M. and Campbell, K. W. (2004). The vertical-to-horizontal response
spectral ratio and tentative procedures for developing simplified V/H and verticaldesign spectra, J. Earthquake Engineering, Vol. 8, pp. 175-207.
Broekhuizen, D. S. (1996). “Effects of vertical acceleration on prestressed concrete
bridges.” MS Thesis, University of Texas at Austin, Texas.
Button, M.R., Cronin, C.J. and Mayes, R.L. (2002). “Effect of Vertical Motions on
Seismic Response of Bridges.” ASCE Journal of Structural Engineering, Vol.128(12), 1551-1564.

283

CalTrans (2006), “Seismic Design Criteria”, California Department of Transportation
Sacramento, CA.
Collier, C.J. and Elnashai, A.S. (2001). “A Procedure for Combining Vertical andHorizontal Seismic Action Effects.” Journal of Earthquake Engineering, Vol. 5,
No. 4, 521-539.
Gloyd, S. (1997). “Design of ordinary bridges for vertical seismic acceleration.” Proc.,
FHWA/NCEER Workshop on the National Representation of Seismic Ground
Motion for New and Existing Highway Facilities, Tech. Rep. No.
NCEER-97-0010, National Center for Earthquake Engineering Research, State
Univ. of New York at Buffalo, N.Y., pp.277–290.
Kunnath, S.K., Erduran, E., Chai, Y.H. and Yashinsky, M. (2007). “Near-Fault Vertical
Ground Motions on Seismic Response of Highway Overcrossings.” ASCE
Journal of Bridge Engineering, Vol.13, No.3, 282-290.
Papazoglou, A.J. and Elnashai, A.S. (1996). “Analytical and Field Evidence of theDamaging Effect of Vertical Earthquake Ground Motion.” Earthquake
Engineering and Structural Dynamics, Vol. 25, No.10, 1109 – 1137.
Saadeghvaziri, M. A., and Foutch, D. A. (1991). “Dynamic behavior of R/C highway
bridges under the combined effect of vertical and horizontal earthquake
motions.” Earthquake Engineering and Structural Dynamics,Vol 20,
pp.535–549.
Silva, W. J. (1997). “Characteristics of vertical ground motions for applications toengineering design.” Proc., FHWA/NCEER Workshop on the National
Representation of Seismic Ground Motion for New and Existing Highway
Facilities, Tech. Rep. No. NCEER-97-0010, National Center for Earthquake
Engineering Research, State Univ. of New York at Buffalo, N.Y., pp.205–252.
Yu, C-P., Broekhuizen, D. S., Roesset, J. M., Breen, J. E., and Kreger, M. E. (1997).
“Effect of vertical ground motion on bridge deck response.” Proc., Workshop on
Earthquake Engineering Frontiers in Transportation Facilities, Tech. Rep. No.
NCEER-97-0005, National Center for Earthquake Engineering Research, State
Univ. of New York at Buffalo, N.Y., pp.249–263.

284

SEISMIC DESIGN OF MULTI SPAN CONTINUOUS RIGID-FRAME
BRIDGE WITH PRESTRESSED CONCRETE BOX GIRDER―NEW-TOMEI GUNKAI-GAWA BRIDGE―
Chiaki NAGAO1 Yasushi KAMIHIGASHI2 Akio KASUGA3 Kenichi NAKATSUMI4
Abstract
The Gunkai-gawa Bridge is a seven-span continuous rigid frame box girder
bridge made of prestressed reinforced concrete, which will be 740 m long and locatednear the Toyota-higashi Junction of the New Tomei Expressway. It was designed tohave strong earthquake-resisting capacity and simple maintenance, by means ofrigid-coupling construction between all piers and main girders. There are threetechniques: first, lightweight upper structure, second below-grade footings for shortpiers, and the last horizontal stress application for displacement adjustment during
closure of the main girders. This paper presents an overview of each process for theseismic design and the approach to construction of the bridge.
1. Introduction
The Gunkai-gawa Bridge will be located 3 km east of the Toyota-higashi
junction on the New Tomei Expressway. The bridge will be 740 m long, a seven-spancontinuous rigid frame box girder bridge made of prestressed reinforced concrete. The
building construction work of this bridge was ordered including designing andplanning to construction. This ordering system enables us to require structural
feasibility, of course, simple maintenance, consideration for surrounding environment,
and what is more, shortened work schedule. The selected design was the bridgeemphasized simple maintenance having no supports due to achieving a rigid-coupledprestressed reinforced concrete structure between all of the piers and main girders.
Generally it is difficult to have rigid-frame bridge because the height of the bridge piers
are lower in comparison to a span of 740 m long bridge. Here, there are three methods
enabling the structure feasible:(1) Lightweight superstructure(2) Below-grade footings applied to short piers(3) Horizontal stress application method for displacement adjustment during closure of
the main girders
This paper presents an overview of the process and its each issue through which
the structural form of the Gunkai-gawa Bridge was determined. It also treats the1

Engineer, Planning and Design Team Nagoya Branch, Central Nippon Expressway
Company Limited
2
Director, Planning and Design Team Nagoya Branch, Central Nippon Expressway
Company Limited
3
Chief Engineer, Civil Engineering Designing Department, Sumitomo Mitsui
Construction
4
Engineer, Civil Engineering Designing Department, Sumitomo Mitsui Construction

285

approach to construction of the bridge.
2. Overview of Gunkai-gawa Bridge
The control points of the Gunkai-gawa Bridge are the Gunkai-gawa (Gunkai
River) and two prefectural roads, located along each bank of the river. Trying to savethe span crossing these control points 100 m level, a bridge pier would need to beplaced between them and the A1 abutment. Accordingly, the length of the spancrossing the river was made 124 m in order to reduce the number of bridge piers. It wasalso determined that using 100 m spans for the other sections would further reduce thenumber of bridge piers. Consequently, a seven-span structure is confirmed optimal.
Since simple maintenance was also required as a designing performance, concrete
girders were used throughout the construction, and a rigid frame structure was used forall bridge piers in order to reduce the number of bearings.
As a result of the establishment of these constraints and the determination of thebridge pier and abutment positions based on further consideration of economy, simple
maintenance and high construction performance, the bridge structure was finalized asa prestressed concrete seven-span continuous rigid frame box girder bridge measuring
740 m long. (Fig. 1)
The ratio of fixed span length to bridge pier height of this bridge isapproximately 8:1. Considering the ratio in an ordinary rigid frame bridge is 5:1 or less,the bridge piers will be affected a major impact from the expansion and the contractionof the main girders. Therefore, the structure was achieved through the incorporation of
measures such as a lightweight superstructure, below-grade footings for short piers,and horizontal stress application for displacement adjustment prior to closure of thea) Side Viewb) Overall Plan View
Fig.1 General View of Gunkai-gawa Bridge286cantilever erection for the outermost piers.
As shown in Fig. 2, themain girders have a wide sectional
shape, so a single-cell box girder
section with an extended slab
reinforced with struts was adopted.
Moreover, the use of a diagonal
web enabled the main girderweight to be reduced by 18% ascompared to a conventional
single-cell box girder section.
Long sections with the same
girder height were provided andstruts of the same length were used,greatly increasing the ease of
construction. Moreover, to reduce
the number of external tendons
within the narrow box girders,high-strength prestressing
strands were used. The main
girders were designed using thelimit state design method, andconcrete with a design standard
strength of σck = 50 N/mm2 wasused in places to reduce theweight.

Fig.2 Main Girder Cross Section

The design strength of
the bridge pier concrete was σck= 30 N/mm2 (40 N/mm2 in some
sections), and SD 345 was used
for the reinforcements. The
types of foundation used were alarge diameter caisson
foundation for all bridge piers,a spread foundation for the A1
bridge abutment, and a caisson
pile foundation for the A2
abutment. As shown in Fig. 3,
shapes of all bridge piers were
5.0 x 5.0 m hollow section, in
which D51 (SD 345) in twoplanes at a maximum as axial
reinforcements and D29 hoop
ties were provided.

Fig.3 Pier Cross Section287

3. Seismic design
Two stages of ground motion, Level 1 and Level 2, were considered in theseismic design of the bridge. As the bridge was a continuous rigid frame bridge and the
1st vibration mode would be dominant, the static check method was used as the methodof checking earthquake-resisting capacity with respect to a Level 1 ground motion. As
adequate study has not been conducted for the applicability of the equal energy
assumption based on the non-linear hysteretic behavior of the structural members andthe bridge as a whole, the active check method was used for the Level 2 ground motion.
3-1. Analysis model and analysis method
A three-dimensional framework model created by modeling the bridge piersand main girders as bar elements was used as the analysis model. The A1 and A2 ends
are support structures, so spring support was used for the bridge axial direction (Xdirection) and vertical direction (Z direction). As a displacement limit structure was
employed for the perpendicular direction (Y direction), fixed support was used for thisdirection. For the bottom end of each bridge pier, a spring constant was considered forthe large diameter caisson foundation. The analysis took into account material
non-linearity for all members (M-φ model).
3-2. Eigenvalue analysis
Fig. 4 shows the major natural periods and vibration modes for the bridge. For
the bridge axial direction, the 1st mode was dominant, and the natural period was 0.948
seconds. In the direction perpendicular to the bridge axis, the 1st and 2nd modes were
dominant. In each of these modes, the main girder with the longest span length (124

P1

P2

P3

P4

P5

P6

a) Axial Direction (X direction)

1st Mode: T=1.698(s)2nd Mode: T=1.597(s)
P6

P6

P5

P5

P4

P4
P3

P3
P2

P2

P1

P1

b) Perpendicular Direction (Y direction)
Fig.4 Vibration Mode288m) was deformed in the direction perpendicular to the bridge axis. The natural periods
were 1.698 and 1.597 seconds, respectively.
3-3. Check of Level 1 ground motion
Under Level 1 ground motion, the design horizontal seismic coefficient wasdetermined by the static check method. That is, the value was kH=0.2 calculated fromits own natural period T=0.948(s) in the bridge axial direction, and the value was kH=
0.15 with natural period T = 1.698 (s) in the direction perpendicular to the bridge axis.
Table 1 shows the results of the check for the bridge piers for the bridge axial
direction and the direction perpendicular to the bridge axis. In the bridge axial direction,piers P3 and P4 in the center had a lower height than the other bridge piers. The section
force during an earthquake will be concentrated on these lower piers, which will be
subjected to the most intense reinforcement stress.
In the direction perpendicular to the bridge axis, as noted in the former section
about natural period modes, the dominant mode was the one in which the main girderwith the longest span length of 124 m was deformed in the direction perpendicular tothe bridge axis. Hence the bridge pier supporting this span, P2 and P3, will be subjected
to substantial reinforcement stress.
Table1. Pier Check of Level 1 Ground Motion
P1

P2

P3

P4

P5

P6

40N/mm2
30N/mm2
30N/mm2
30N/mm2
30N/mm2
40N/mm2
SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150
Axial Direction
2 line2 line2 line2 line2 line1.5 line
SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150 SD345 D51@150
perpendicular direcion
2 line2 line1 line1 line2 line2 lineDesign Strength of Concrete
Axial
Reinforcing
Bar

Axial Direction

D29@150

D29@150

D29@125

D29@150

D29@150

D29@150perpendicular direcion

D29@150

D29@150

D29@125

D29@150

D29@150

D29@150

13.2
19.5
O.K.
214

11.9

14.1

13.3

11.8

O.K.
270

O.K.
194

13.7
19.5
O.K.
228

O.K.
8.8

O.K.
11.0

O.K.
142

O.K.
171

O.K.
10.5
19.5
O.K.
145

O.K.

O.K.

O.K.

Hoop tie2σc(N/mm )σa(N/mm2)
Axial
Direcionσs(N/mm2)
σa(N/mm2)
σc(N/mm2)
σa(N/mm2)

Perpendicular
Direcionσs(N/mm2)
σa(N/mm2)

15.0
O.K.
189

O.K.
284
300

O.K.
11.7
19.5
O.K.
177

O.K.
11.3

O.K.
10.2

O.K.
172

O.K.
176

O.K.

O.K.

O.K.

15.0

300

σc:Concrete Stress
σs:Reinforcing Bar Stress
σa:Limit

3-4. Check of Level 2 ground motion
The required earthquake-resisting capacity with respect to a Level 2 groundmotion was based on Specifications for Highway Bridges (Part V: Seismic Design).
The seismic design was conducted so as to satisfy earthquake-resisting capacity 2(which means to be in limited damage resulting from an earthquake Level 2 which canbe restored temporary bridge performance quickly). The input ground motion used for289

the check was an acceleration waveform based on Specifications for Highway Bridges(Part V: Seismic Design).
(1) Check of bridge piers
The bridge piers were figured out to have enough performance by checking inthree items in the bridge axial direction and the direction perpendicular to the bridge
axis: (1) maximum response curvature at top and bottom ends (2) shear capacity, and(3) residual displacement. The results are shown in Table 2. The values for each check
item satisfied the requirements for allowable limits.
Table2. Pier Check of Level 2 Ground Motion
Unit

P1

P2

P3

P4

P5

P6

response
limit
ratio

(rad)
(rad)

response
limit
ratio

(rad)
(rad)

response
limit
ratio

(rad)
(rad)

response
limit
ratio

(rad)
(rad)

response
capacity
ratio

(kN)
(kN)

response
capacity
ratio

(kN)
(kN)

response
capacity
ratio

(kN)
(kN)

response
capacity
ratio

(kN)
(kN)

TypeⅠ

response
limit

(m)
(m)

TypeⅡ

response
limit

(m)
(m)

TypeⅠ

response
limit

(m)
(m)

TypeⅡ

response
limit

(m)
(m)

1.5E-03
4.4E-0335%
OK
2.0E-03
1.3E-0216%
OK
3.1E-03
4.5E-0368%OK
1.5E-03
1.3E-0211%OK
2.7E+04
3.7E+0472%
OK
3.1E+04
3.8E+0481%
OK
1.9E+04
3.7E+0451%
OK
1.5E+04
3.8E+04
40%
OK
0.033
0.400
OK
0.052
0.400
OK
0.060
0.400
OK
0.000
0.400
OK

1.7E-03
4.3E-0339%
OK
2.0E-03
1.2E-0216%
OK
3.8E-03
4.4E-0387%
OK
1.7E-03
1.3E-0213%
OK
2.6E+04
3.7E+0471%
OK
2.8E+04
3.8E+0474%
OK
1.8E+04
3.7E+0447%
OK
1.5E+04
3.8E+04
40%
OK
0.030
0.432
OK
0.049
0.432
OK
0.064
0.432
OK
0.000
0.432
OK

3.7E-03
4.9E-0376%
OK
6.2E-03
1.7E-0236%
OK
1.6E-03
4.2E-0338%
OK
1.4E-03
1.3E-0211%OK
3.6E+04
4.4E+0480%
OK
3.8E+04
4.5E+0484%
OK
1.7E+04
4.4E+0439%
OK
1.6E+04
4.5E+04
35%
OK
0.026
0.267
OK
0.044
0.267
OK
0.015
0.267
OK
0.003
0.267
OK

4.0E-03
4.8E-0384%OK
5.9E-03
1.6E-0236%
OK
1.7E-03
4.4E-0338%
OK
1.5E-03
1.4E-0211%
OK
3.4E+04
3.7E+0492%
OK
3.6E+04
3.8E+0496%OK
1.8E+04
3.7E+0448%
OK
1.8E+04
3.8E+04
48%
OK
0.025
0.267
OK
0.041
0.267
OK
0.016
0.267
OK
0.010
0.267
OK

2.0E-03
4.4E-0347%
OK
2.7E-03
1.3E-0221%
OK
4.3E-03
4.5E-0395%
OK
2.1E-03
1.3E-0216%
OK
3.4E+04
3.7E+0492%
OK
3.6E+04
3.8E+0495%
OK
2.1E+04
3.7E+0456%
OK
2.0E+04
3.8E+04
53%
OK
0.026
0.347
OK
0.042
0.347
OK
0.082
0.347
OK
0.020
0.347
OK

1.5E-03
4.0E-0337%
OK
1.7E-03
1.0E-0216%
OK
3.3E-03
4.1E-0382%
OK
1.6E-03
1.1E-0215%
OK
2.5E+04
3.0E+0483%
OK
2.7E+04
3.0E+0489%OK
1.9E+04
3.0E+0463%OK
1.7E+04
3.0E+04
55%
OK
0.028
0.425
OK
0.044
0.425
OK
0.063
0.425
OK
0.000
0.425
OK

(1)Maximum Respons Curvature

TypeⅠ
Axial Direcion
TypeⅡ

TypeⅠ
Perpendicular
Direcion
TypeⅡ

TypeⅠ

(2)Shesr Capacity

Axial Direcion
TypeⅡ

TypeⅠ
Perpendicular
Direcion(3)Residual Displacement

TypeⅡ

Axial Direcion

Perpendicular
Direcion(2) Check of main girders
In the check conducted for the bridge axial direction, it was confirmed that the290response bending moment in the in-plane direction acting on the main girder was
within the initial yield moment of the axial direction reinforcements in both the top of
the upper slab and the bottom of the lower slab (Fig. 5). To resist positive bending, theaxial direction reinforcements in the bottom slab were upgraded throughout the entire
length of the bridge, which were replaced up to D25 at 125 mm pitch. In addition,negative bending was increased at side spans according to vibration mode in Fig.4 a).
That is due to the great length of the side spans at both ends (92.8 m) and the fact thatthe ends were supported by bearings. So the axial direction reinforcements for the top
slab near the center of the span were also upgraded to D25 at intervals of 125 mm.
The check conducted for the direction perpendicular to the bridge axis
confirmed that the response bending moment in the out-of-plane direction acting on themain girder was within the initial yield moment of the axial direction reinforcementson the outside of the web.
-1000.0

Respons(TypeⅠ)-900.0

Respons(TypeⅡ)-800.0Initial Yield-700.0-600.0-500.0curvature(1/m)-400.0-300.0-200.0-100.0
0.0
100.0
200.0
300.0
400.0
500.0
600.0

A1

P1

P2

P3

P4

P5

P6

A2

Fig.5 Maximum Response Curvature on the Main Girderof Level 2 Ground Motion

4. Approach to future construction
On this bridge, the eastbound and westbound lines are separated, but as accessis restricted in some areas, the eastbound lanes will be constructed first. The major
overall processes for the construction are as follows:
1. Ground leveling for construction road and work yard
2. Caisson foundation construction
3. Bridge pier construction
4. Pier caps291

5. Cantilever erection work
6. Bridge surfacing
The cantilever erections of the main girders will be built up at once for all
bridge piers to shorten the construction time.
Below is a description of the below-grade footings for short piers and thehorizontal stress application method for displacement adjustment during closure of themain girders, methods that were adopted to achieve the characteristic rigid frame
structure of the bridge.
(1) Below-grade footings for short piers
Below-grade footings are used for short piers to achieve the continuous rigid
frame structure. On lower piers P3 and P4, the top surfaces of the foundations areplaced below the surface of the ground to adjust the pier height. There are kept hollow
with covers without backfilling between these piers and ground to prevent the pierfrom being subjected to ground resistance in the event of an earthquake. The cover canbe removed and steps are hung to enables us to descent to the base of the pier for
inspection after an earthquake. (see Fig. 6).

Fig.6 Below-Grade Footings for Short Piers(2) Horizontal stress application for displacement adjustment
As this bridge has a long fixed span, the creep and dry shrinkage of concretethat may be large after the main girders have been completed will increase the bending
moment applied to the end piers. Normally, under the condition of the bridge being
subjected to dead load and temperature fluctuations, the section at the base of the end
piers must be larger than those of the other bridge piers. As a result, the foundation
structure of the end piers must also be larger.
In order to avoid this, before the main girders between end piers P1, P6 and thenext pier P2, P5 are closed, horizontal force is applied to reduce the bending moment292

that acts on the end piers. It enables to reduce the section force about 28% and to keep
the reinforcement stress within the acceptable limit. Fig. 7 shows a conceptual diagramof the adjustment method.

Fig.7 Horizontal Stress Application
5. Conclusion
This paper has described the structural form adopted for the Gunkai-gawaBridge, which was ordered as a design and build package project. There were threemethods used to achieve the structure of this bridge. An overview of the seismic designof the structure was also presented. There are few examples of the construction of long,continuous rigid frame structures such as this bridge that also has a low bridge pierheight. It was confirmed that the bridge which has a rigid-coupled structure thatemphasizes simple maintenance also provides excellent earthquake-resisting capacity.
As of September 2009, construction of the Gunkai-gawa Bridge was at thestage of leveling the land for the construction road, so the work has just begun. A
detailed construction plan will be drafted for the completion of the bridge, andthorough attention will be given to quality control and safety management.

293blank25th US-Japan
Bridge Engineering Workshop

Session

7

Seismic Retrofit
Seismic Retrofit Study of CABLE-STAYED Bridge on TOKYO-GAIKAN EXPRESSWAY
By Tsutomu Yoshioka, Yoshinori Kawahira, and Kouichirou Shitou
Repair of High Shear Standard Reinforced Concrete Bridge Columns Using Cfrp
By M. "Saiid" Saiidi and Ashkan Vosooghi
Seismic Retrofit Techniques for Reinforced Concrete Bridge Columns with Combination of
FRP Sheet and Steel Jacketing
By Guangfeng Zhang, Shigeki Unjoh, Jun-ichi Hoshikum, and Junichi Sakai
Development and Refinement of Illinois' Earthquake Resisting System Strategy
By Daniel H. Tobias, Jerome F. Hajjar, Ralph E. Anderson,
James M. LaFave, Chad E. Hodel, Larry A. Fahnestock, William M. Kramer,
Joshua S. Steelman, Patrik D. Claussen, Kevin L. Riechers, and Mark D. ShafferPage Left Blank

SEISMIC RETROFIT STUDY OF CABLE-STAYED BRIDGEON TOKYO-GAIKAN EXPRESSWAY
Yoshinori Kawahira 1, Kouichirou Shitou2 and Tsutomu Yoshioka3
Abstract
This paper describes the seismic performance verification and retrofit method
examination of a cable-stayed bridge in the Sakitama Bridge. First, the input
earthquake motion was specified for use in both the target seismic performance and thebridge verification, while the parts and members of the latter were verified by seismic
response analysis. The main tower and caisson foundation that were difficult toevaluate were verified by a nonlinear finite element analysis. Based on the verification
results, regions requiring a seismic retrofit were identified, and the retrofit methods
were examined.
Introduction
The imminence of large-scale earthquakes, namely the Tokai, Tonankai andNankai Earthquakes has been pointed out in Japan in the near future. During the threeyear period from FY 2005 to FY 2007, the Ministry of Land, Infrastructure, Transport
and Tourism has focused on the seismic retrofit of bridges on emergency routes thatwould play important roles during earthquakes [1]. Under this Three-Year Program, in
addition to bridges built before the 1980 edition of the Design Specifications for
Highway Bridges was established, special structures and long-span bridges were also
designated as targets of the seismic retrofit.
As the emergency transportation route, the Tokyo-Gaikan Expressway would
assume vital roles in rescue and relief activities and emergency transportation during
a large-scale earthquake (Fig. 1). The Sakitama Bridge on this Tokyo-Gaikan
Expressway is located in Bijogi, Toda City, Saitama Prefecture, and crosses the
Arakawa River (Fig. 2). The cable-stayed bridge in Sakitama Bridge has a special
structure, and was therefore selected for the examination of a seismic retrofit under thisThree-Year Program. This paper describes both the seismic performance verification
and retrofit method of the cable-stayed bridge in Sakitama Bridge.
The Outline of the Studied Bridge
Figure 3 shows a general view of a cable-stayed bridge in the Sakitama Bridge.
This cable-stayed bridge is carrying both the expressway and general road (National
Highway Route 298) of the Tokyo-Gaikan Expressway. The two bridges are arranged
in a parallel for inbound and outbound routes.
This cable-stayed bridge lays the cables in a multi-fan type in a single plane. It
1Head Manager Management Section, Kitashuto National Highway Office
Deputy director, Misato management office, East Nippon Expressway Co.,Ltd.
3
Chief, Maintenance Engineering Research Institute, Nippon Engineering Consultants
Co., Ltd.
2297

Fig.1 Location of Sakitama Bridge

Fig.2 Photo of Sakitama Bridge

Fig.3 General view of Sakitama Bridge
is 2-span continuous steel cable-stayed bridge with a span length of 190m. The main
tower is a single post type and made of steel. The main girder is a steel box girder with
3-cell, with a total width of 26m. The main tower and main girder are rigidly connected,
and connected to the pier via the bearing support. The piers from P11 to P13 are all of
the RC wall type. The bearing support condition in the longitudinal direction is fixed
only on P12 pier, the middle support. A pivot bearing is installed on P12 and pivot
roller bearings are installed on P11 and P13, which are the end supports. The bearing
support condition in the transverse direction is all fixed, and wind bearings to be used298

for fixing the transverse direction are installed on P11 and P13, which are the edge
supports. Rocking bearings for controlling uplift are also installed on P11 and P 13.
The ground is predominantly soft to a depth of about 25m from the surface; mainlycomprising a cohesive soil layer, the N value of which is around 3 to 11. The primarynatural period of the surface ground is 0.68 to 1.08 sec., and the ground is specified as
Ground Type III in the Design Specifications for Highway Bridges. The foundation
type of P11 is a steel-pipe-sheet pile foundation, and those of P12 and P13 arepneumatic caisson foundations.
This bridge is based on the design standard specified in the Design
Specifications for Highway Bridges of 1980 that was issued before the seismic design
standards on highway bridges were revised in 1996, following the Hyogo-ken Nanbu
Earthquake in 1995. Therefore, this bridge is designed to ensure seismic performance
only for earthquakes (level 1 earthquake motion) highly likely to occur during theservice life, and may not potentially satisfy the seismic performance for earthquakes(level 2 earthquake motion) not likely to occur but of high intensity. The cable-stayed
bridge for the inbound route has been in service since 1992, and that for the outboundroute since 1998.
Seismic Retrofit Basic Policy
Based on the Seismic Design Edition of the Design Specifications for Highway
Bridges [2] and the Three-Year Program of Seismic Retrofit of Bridges on Emergency
Routes [1], the target seismic performance for the level 2 earthquake motion incurredby the cable-stayed bridge in Sakitama Bridge requires that it should sustain only
limited damage from level 2 earthquake motion, and be capable of swiftly reverting to
its main function as an emergency road. The performance is intended to ensure safety
against the risk of the bridge collapsing, and facilitate the traffic of emergency vehicles
for disaster recovery activities after earthquakes. The performance is also intended toallow ordinary vehicles to use the road only with emergency repairs, and allow
permanent repairs to be made while the bridge is in service.
To meet the above-mentioned requirements of the target seismic performance,
the critical state of the structural elements comprising the bridge must be set. Table 1
shows the division of the damage level of each member and the required performancefor level 2 earthquake motion. The main girder is intended to avoid local buckling on
the deck plates in the driving lanes of emergency vehicles, and the floor system that
supports the deck plates is intended to yield only slightly and remain strong enough tobear the load of the emergency traffic. The main tower is intended not to succumb tolocal buckling, but to yield slightly in its marginal section, and remain strong enough tobear the load of emergency traffic. The cables are intended neigher to succumb tofalling away due to loss of cable tension, nor break when subject to tensile force. The
bearing is intended to maintain the force applied to the component parts below the
maximum strength. The pier is intended to tolerate major ductility but to confine it toan extent facilitating prompt repairs. The foundation is intended to tolerate ductility
provided the ductility limits deformation to a level that is not destructive to the overall
safety of the bridge system.

299

Table.1 Classification of damage level of bridge members for level 2 earthquake
Degree of damage for structure
Bridge members
In planederection

Main
girderSlight damage
Damage level I

Damage level II

Severe damage
Damage level III

A state that occurrence stress degreeslightly exceeded a yield stressdegreeA state that the deck plate which
stretched caused local buckling, andthe box girder inside slightly yieldA state in the domain that was stable A state that a strength as the main
as for the strength and the ductility beam system has begun to deteriorate
although the plasticity spread
A state that local buckling andA state that a strength as the main
plasticity progressed to the box girder beam system has begun to deteriorate
insideA state that occurrence stress degreeslightly exceeded a yield stressdegree, But local buckling at theelastic level does not occur
A state that tensile stress degreeexceeded a yield stress degree (0.7%
elongation strength) in one cable

A state in the domain that was stable A state that a strength as the main
as for the strength and the ductility tower system has begun to deteriorate
although the plasticity spread

Out plane
derection

Main tower

stay cableσ > σ y ( = σ u 1.4 )Bearing support

A state that tensile stress degree
A state that a cable was falling away
exceeded a yield stress degree (0.7% by tension loss in plural cables, andelongation strength) in plural cables tensile stress degree exceeded breakstrength
σ > σuσ > σ y ( = σ u 1.4 )A state that occurrence horizontal
force exceeded the smallest yieldstrength in their component parts

A state that occurrence horizontal
force exceeded the smallest
maximum strength in theircomponent parts
P>P

Py＜ P≦ Pu
RC pieru

A state that cracking of the concrete
and yield of the reinforcement occur,
but have still redundancy for
maximum strength

A state that cracking of the concrete
and yield of the reinforcement
spread, but have stability for thestrength and the ductility

A state that a strength as the piersystem has begun to deteriorate

1.0＜ μ r ≦ μ a

μ r > μ a S > Ps δ R > δ Ra

A state that plasticity of the aroundground and foundation slightly float

A state that foundation yield, andfoundation floating and plasticity of
the around ground spread throughA state that foundation strength
deteriorated and lost stability of thesuperstructure

1.0＜ μ r ≦1.2

Foundation(P12,13 : caisson)(P11 : Steel pipe sheet
pile)* The bold frames show that it is a limit state to satisfy required performance.

Figure 4 shows an outline of the seismic retrofit examination on the cablestayed bridge. After specifying the target seismic performance for level 2 earthquake
motion, the seismic performance verification of the current structural system isperformed via seismic response analysis. Regarding the input earthquake motion, we
initially examined the need to prepare site waves taking regional characteristics into
consideration. Consequently, it emerged that there was no information concerning theactive fault near the Sakitama Bridge that required realistic consideration of anearthquake. Therefore, we decided to use the standard waves (type II of level 2
earthquake motion) in the Design Specifications for Highway Bridges establishedbased on the Hyogo-ken Nanbu Earthquake of 1995. For plate-boundary earthquakes,
we also decided to use the standard waves (type I of level 2 earthquake motion)in the
Design Specifications for Highway Bridges because we considered that this bridge
would be affected mainly by an inland direct strike type earthquake, and that no effect300

Fig.4 Flow of seismic retrofit design
Acceleration wave I-III-1

Type2-III

0
Maximum； 3.714（40.33s）
Minimum；-4.334（41.38s）

-5

-10

1000

0

10

20

30
time(sec)

40

50

60

(a) Type1 ground motion
10

2

Acceleration （m/s ）Type1-III

5

Acceleration response spectru （gal）

Acceleration （m/s ）

2

10000

10

Acceleration wave II-III-1

50

P12-Transverse
1st mode 0.35s

100

P12-Longitudinal
1st mode 0.88s

Maximum； 4.653（ 6.79s）
Minimum；-5.910（ 6.30s）

-5

-10
0

10

20

30
time(sec)

40

50

60

Tower-Out plane
1st mode 2.67s

Girder-Lateral bending
2nd mode 0.90s

1001Natural period （sec）(b) Type2 ground motion(c) Acceleration response spectrum
Fig.5 Design earthquake ground motion
could be expected from the preparation of site waves. Figure 5 shows the time history
response waveform and acceleration response spectrum of the representative 1 wave oftypes I and II.
Seismic Response Analysis Model
Before performing a seismic response analysis, we calculated the lateralstrength of the bearing support, pier and foundation, and estimated the damage
sequence. As shown in Fig. 6, the lateral strength of the caisson foundation of P13 in

301

10

Lateral Strength (kN)

100000

Bearing support
Pier-2nd cut-off reinforcement
Pier-1st cut-off reinforcement
Pier-bottom of column
Foundation

120000100000Lateral Strength (kN)

120000

80000
60000
40000
20000

Bearing support
Pier-2nd cut-off reinforcement
Pier-1st cut-off reinforcement
Pier-bottom of column
Foundation

80000
60000
40000
20000

0

0
P11(M)

P12(F)

P13(M)

P11(F)

P12(F)

P13(F)(a) Longitudinal direction(b) Transverse direction
Fig.6 Comparison of lateral strength capacity of each members

Fig.7 Analytical model
the transverse direction is lower than that of the pier, and the surface ground is soft.
Accordingly, it is necessary to perform seismic performance verification of each partand member via seismic response analysis that takes account of the ductility of thefoundation. As for the main girder made of orthotropic steel deck 1-box girder, theoverhang of the deck is long, and number of ribs of the sidewalk is small. It is therefore
highly probably that the main girder will buckle by an earthquake in the transverse
direction, and it is advisable to use fiber elements that allow us to properly consider the
nonlinearity of the material in complicated sectional form. However, the calculationload of the analysis that takes account of the plasticity of both the foundation and main
girder is significant, and the calculation result is complicated. As shown in Fig. 7, wetherefore decided to perform seismic response analyses on two cases; one for the casewhere the foundation is assumed to cause no ductility (Model A), and the other casewhere the ductility of the foundation is taken into consideration (Model B).
In Model A, the foundation is replaced with a sway-rocking spring, and themain girder is modeled with a fiber element; likewise the main tower and pier. Each302

cable is subdivided into 8 parts, and mass allocated to take into account the vibrationof cable itself. The rocking bearing is modeled with a truss element, and resistance inthe longitudinal direction is taken into account. To take ductility of the foundation in
Model B into consideration, the caisson foundation is a nonlinear girder element M –φmodel, and ground resistance is modeled with a nonlinear spring. The main tower andpier are also M –φ models and thus the same as the foundation. To restore their force
characteristics, the main tower is a movement hardening type tri-liner model, and thepier and caisson foundation are degrading type tri-liner models, while the main girder
is an elastic beam element. The movable bearings on P11 and P13 are nonlinear springs
that take account of both frictional resistance and movement restriction.
The material and geometrical nonlinear dynamic analysis of these Models A
and B was performed using input seismic motion shown in Fig. 5. Rayleigh damping isused for viscous damping in a dynamic analysis. Rayleigh damping is set to include
main modes by the natural vibration analysis performed in advance (Fig. 8). The
Newmark-β (β=0.25) method is used as the numerical integration. The computation
time interval is 0.002 seconds.
The main girder, and the cable significantly affected by the ductility of the main
girder are verified using the result of Model A, while other parts are verified using thatof Model B.
0.20
NumberThn=2 : 2.666s(0.375Hz)
st(a) Tower - Out plane 1 mode

Modal damping constant

0.18
0.16

2

43
0.134
0.079
β
0.00335

2.666
0.016
α
0.05468

0.14
0.12
0.10
0.08
0.06
0.04
0.02
0.00
0

1

2

3

4

5

6

7

8

9

10

Natural frequency (Hz)n=8 : 0.900s(1.111Hz)(c) Modal damping and Rayleigh damping
model applied (Transverse direction)(b) Main girder – Lateral bending 2nd mode
Fig.8 Natural vibration modes and modal damping (Model A)
Seismic Performance Verification of Members
This section describes the details of the seismic performance verification
performed using the seismic response that was calculated by material and geometricalnonlinear analysis, and also using either the load-carrying capacity or deformation
performance of the bridge members.
The main girder was significantly affected by the earthquake in the transverse
direction. The result of Model A that used fiber elements showed local ductility in boththe sidewalk deck plate and the box girder, up to about 4εy, and 2.5εy respectively (Fig.
9). However, the nonlinear finite element analysis results suggested that the undulation303

of the road surface after local bucking was so small that the serviceability of the road
could be ensured after earthquake (Fig. 10). Consequently, the damage level was
evaluated as I, and no retrofit was deemed necessary.
Max Vertical
Displacement=about 32mm

3εy2εy-3εyεy-2εy

P11

P12

-εy

Deck Plate

Fig.9 Strain distribution of deck plate(Dynamic analysis result by Model A)

Fig.10 Deformation of deck plate(Nonlinear FEM analysis result)

The cables were affected predominantly by an earthquake in the longitudinal
direction, however, the maximum tensile stress caused by earthquake was 773N/mm2,
namely well below the 0.7% elongation strength σy (=1160N/mm2) (Fig. 11). The
minimum tensile stress was 151 N/mm2 , and the cable has no tension loss. Accordingly,
there is no damage.
1600

Break strength σu

1400

Response curvature φr
70

0.7% elongation strength σy60

Allowable stress σa

50

Altitude (m)Stress (N/mm 2）.

1200

80

： M inimum by earthquake： By dead load
： M aximum by earthquake

1000
800
600

40
30

400

20

200

10

0
1st

2nd

3rd

4th

5th

6th

7th

Fig.11 Performance verification result ofstay cables (Longitudinal direction,
Type2 earthquakes)

Yield curvature ofcompressive edge φyc
Yield curvature oftens ile edge φyt

0
0.000

0.001

0.002

0.003

Curvature (1/m)

Fig.12 Curvature distribution of tower(Transverse direction, Type1 earthquakes)

The main tower was affected by a type I Earthquake in the transverse direction
such that it caused local buckling in the elastic region within the range of about 26mfrom the foundation. For this reason, longitudinal ribs are added to control thewidth-thickness ratio parameter and keep it within the limiting value specified in the
Design Specifications for Highway Bridges. Assuming that the parameter was304improved, the post-yielding behavior was observed, whereupon the tower was seen toshow slight ductility within a range of about 8m from the foundation (response
plasticity rate μr=1.38) as shown in Fig. 12. The P-δ curve of the main tower was
calculated by nonlinear finite element analysis. The results showed that the towerlacked toughness of the SM570 high strength steel, and that its ductility capacity wasevaluated as around 1.79 (Fig. 13). It was so close to the response value and difficult toevaluate the damage as level I, consequently, ductility enhancement measures weredeemed necessary to reduce the deformation.
Table 2 shows the result of verification of the bearing supports. The maximum
response horizontal force of the pivot bearing on P12 exceeded the yield strength,
however, it was lower than the maximum strength. Consequently, the damage level
was evaluated as level I, and the pivot bearing was deemed to satisfy the required
performance. The maximum horizontal displacement of the pivot roller bearings on
P11 and P13 exceeded the movement capacity, and the upper bearing collided with the
stopper. However, the maximum horizontal force generated on the stopper was lessthan the maximum strength, hence no retrofit was deemed necessary. The maximum
tensile force generated on the rocking bearing was less than the yield strength, hence
the rocking bearing was to cause no damage. The maximum horizontal force generated
on the wind bearing exceeded the maximum strength, hence suggest the potential for
μa=1.79
μr=1.386000

μr／μa=0.77
Maximum point
3.568

Maximum response(Dynamic analysis)

2.747

Horizontal force P
(kN)

5000
Yield point

1.993

4000
3000
2000
13710.

When tower box was
retrofited by ribs

1000

828883.

0
0.0
0.5
1.0
1.5
2.0
2.5 3.0
3.5
Residual displacement 27cm Horizontal displacement δ (m)

4.0

4.5

5.0

Fig.13 Nonlinear FEM analysis result of tower (Transverse direction,
Type1 earthquakes)
Table.2 Performance verification result of bearing supports
Bearing support

SupportNo.

Responsequantity

Pivot bearing

P12

Force (kN)

Pivot roller
bearing

Rocking bearing
Wind bearing

P11
P13
P11
P13
P11
P13

Displacement (mm)Force (kN)
Displacement (mm)Force (kN)
Tension (kN)
Force (kN)

Seismic
direction
LG
TR
LG
LG
LG
LG
LG
LG
TR
TR

305

Maximum
response
56737
53935
110
2565
106
0
14476
6035
19822
17621

Capacity
Yield
Maximum
strength
strength
50460

85782

110
1512
110
1512

230
2570
230
2570

20937

27561

11520

19584

Damage
level
I
I
I
0
0
0
III
III

the components to be broken and moved laterally. If the main girder slides laterally,loads carried by the rocking bearing sharply increases, thus causing the rocking bearing
to break. Then, uplift of the main girder may be occurred, and the bridge would not be
used for emergency vehicles. For these reasons, the damage level of the wind bearing
was evaluated as level III, and a retrofit was deemed necessary.
Figure 14 shows the maximum response force distribution of the pier that
corresponds to its failure mode. This figure shows the response distribution of bothModels A and B together with that of the foundation. In an earthquake in thelongitudinal direction, piers P 11 and P 13 caused ductility to their bases solely by theinertia force of their own weight. In an earthquake in the transverse direction, themaximum shear force of the entire pier, including the cut-off of longitudinal
reinforcement at mid-height, exceeded the shear strength, hence the high potential forthe piers to succumb to shear failure. Meanwhile, P12 experienced shear failure modein both longitudinal and transverse directions, with the shear force exceeded its
P11
Bending failure mode00Level crown of pierModel A
Model B
Initial yield moment-10

Level crown of pierModel A
Model B

-20
Level crown offoundation-30

Altitude (m)Altitude (m)Level crown offoundation

-30

-50

-50
Bottom of foundation

Bottom of foundation-60

Bottom of foundation-60
100000 200000 300000 400000 500000

-60
0

50000

100000150000200000

0

Shear force (kN)

Bending moment (kN･m)Shear failure after bending
yielding
0

Bending moment (kN･m)

Model A
Model B
Shear strength

Level crown of pierModel A
Model B
Shear strength Ps0

-10

-20

0

Level crown of pierLevel crown of pier-20
Level crown offoundation

-30

-40

-40

-50

-50

20000

40000

Shear force (kN)

60000

-20%

-50
Bottom of foundation-60
0Level crown offoundation

-30

-40

Bottom of foundation-60

Altitude (m)Altitude (m)Level crown offoundation

Model A
Model B
Shear strength

-10

-20

-30

100000 200000 300000 400000 500000

Shear failure after bending
yielding

Shear failure mode

0-10Level crown offoundation

-30

-40

-40

0

Model A
Model B
Initial yield moment

-10

-20

-40

Altitude (m)0Shear strength Ps0-50

Transverse direction

P13
Bending failure mode

Level crown of pier

-10

-20
Altitude (m)Longitudinal direction

P12
Shear failure mode

Bottom of foundation

-60
0

50000

100000150000

Shear force (kN)

2000000

20000

Fig.14 Maximum section force distribution of pier (Type2 earthquakes)

306

40000

Shear force (kN)

60000

maximum shear capacity in both directions. Based on these results, all piers were
evaluated as level III, and some retrofits were deemed necessary. As shown in the shear
force distribution in the transverse direction of P13, the maximum horizontal force
acting on the crest of the foundation diminished by around 20% in cases where theductility of the foundation was taken into account. Consequently, such latter cases
were considered to be closer to the actual behavior. It was decided to use the value of
this case as the acting force to be used to verify the foundation.
Table 3 shows the verification result of the caisson foundation. The verification
was performed by a lateral strength approach; using the acting force of the crest of thefoundation obtained by a dynamic analysis. The verification of the caisson foundationof P12 showed that the shear force of its members exceeded the shear capacity,although satisfying the stability verification as the foundation. During the verificationof the caisson foundation of P13, the shear force exceeded the shear capacity, and its
ductility demand of about 27.0 substantially exceeded the ductility capacity of 5.9. The
circumferential rebar of the sidewall of the caisson foundation of P13 is the main
reinforcement, and bars are arranged using D19@300, while the reinforcement
D16@300 in the longitudinal direction is very low. The verification of hollow RC
section with low reinforcement based on the beam theory is simple, however, thisverification may become over conservative. Therefore, the verification was repeatedusing nonlinear finite element analysis (see the subsequent section). Meanwhile, thesteel-pipe-sheet-pile function of P11 satisfied both the stability and member
verifications.
Table.3 Performance verification result of caisson foundations (Transverse direction,
Type2 earthquakes)
P12 caisson

Performance verification item

Response
Ductility demand of foundation
3.022
54.0
Stability Plasticity domain rate of foundation front ％verification Float area rate of foundation bottom
35.8％
Rotation angle of level crown of foundation rad
0.012
Vertical section of side wall
Shear
kN
77912
Verification
Bending kN･m1277
of members Horizontal section of side wall
Shear
kN
1153

＜
＜
＜
＜
＜
＞
＜

P13 caisson

Capacity Response
Capacity
6.000
26.956 ＞
5.900
60.0
55.8 ＜
60.0
60.0
0.0 ＜
60.0
0.020
0.039 ＞
0.020
16283752817 ＞
28727
1239
1709 ＞
1212
1331
1153 ＞
411

Figure 15 shows a summary of the verification results showing both theevaluation results of the damage levels of this bridge, and the locations where therequired performance were not met when exposed to an earthquake in the transverse
direction
Seismic Performance Re-verification of the Caisson Foundation
Figure 16 (a) shows the FE model of the caisson foundation of P13 that wasverified in detail using nonlinear finite element analysis. On the FE model, the caisson
foundation was modeled with a solid element, and the reinforcement was modeled witha rebar element. The surrounding soil was modeled with a nonlinear spring element,
which was arranged at each joint on the peripheral surface. To reduce the computation,

307

: The box letter shows the part that does not satisfy required performance.
: Damage level I

Main girder :It became plasticity, but serviceability
were satisfied.

: Damage level II: Damage level III

Main tower :Local buckling occurredWind bearing :Exceeded strength
F

△

△

F
Cut-off of longitudinal reinforcement▽: Shear failure

Wind bearing :Exceeded strength
F△ F

F
Cut-off of longitudinal reinforcement: Shear failure

Cut-off part
: Shear failure

Bottom of column : Shear failure

Bottom of column : Shear failure

Bottom : Shear failure
Foundation :Verified in details

Foundation :It yielded, but stability were satisfied.

Foundation :It yielded, but stability were satisfied.

P12

P11

P13

Fig.15 Summary of performance verification result (Transverse direction)

50000

40000
Horizontal force (kN)Maximum strength38250kN

Dynamic response
approximation
33750kN

45000

35000
30000
25000
20000
15000
10000

FEM analysis result

5000

Dynamic analysis result

0
Y

0

Z

X

200

400

600

800

1000

1200

1400

Horizontal displacement of level crown of foundation(mm)(b) P – δcurve(a) FE model
0.000926

Over 3500μ-0.00233(Concrete collapsed)-0.00559-0.00885-0.01210.

-0.0154-0.0035-0.0186-0.0219-0.1-0.0251-0.0284-0.0317
Z

Y
X
Output Set: Step 16, Inc 1377
Deformed(0.491): Total Translation
Contour: Solid Min Prin Strain

Z
Y
X
Output Set: Step 16, Inc 1377
Deformed(0.491): Total Translation
Contour: Solid Min Prin Strain-0.0349-0.0382(c) Minimum principle strain contour
Fig.16 Nonlinear FEM analysis result of P13 caisson (Transverse direction,
Type2 earthquakes)308

a half model was used for the longitudinal direction. The model used in the analysis
involved a 5m portion of the pier and the rigid surface on top of the pier, and the force
obtained from the dynamic analysis was applied gradually to the center. The analysis
software used for this analysis is the general-purpose finite element program ABAQUS
Ver. 6.5.
Figure 16 (b) shows the P-δ curve obtained from the analysis. Stability
verification was performed at the loading step corresponding to the dynamic analysis
response of approximately 200mm, the horizontal displacement of the foundation crest.
The result showed that the ductility demand of 6.4, although slightly exceeding theductility capacity of 5.9 that included a safety factor of 1.8, did not reach the ultimate
ductility factor of 9.8. The dynamic analysis response value was on the upward
gradient region, and assuming that the horizontal strength of 38,250kN wereconsidered, there is 13% margin remained until the strength. Therefore, although thefoundation caused plasticity, it was considered the safety was maintained.
The members were checked for the loading steps that corresponded to the
dynamic analysis response value. As shown in Fig. 16 (c), stress concentration
associated with bending in the upper corner of the sidewall occurred. The right figure
shows the minimum principal strain of the concrete expressed with the contour in twocolors, when the ultimate strain was 3,500με. This stress concentration exceeded theultimate strain, and there was a possibility of causing spalling of cover concrete, andbuckling and expansion of reinforcement. However, the dead load analysis performed
on a model with no members in the circular area of the sidewall revealed that thestresses on the linear section of the side and partition walls were approximately half theallowable stress. Based on this result, it was considered that the members were able tosupport the dead load caused by the level 2 earthquake motion.
As described above, the caisson foundation was reevaluated by nonlinear finite
element analysis, and the result showed that it was able to satisfy the stability
requirements. However, the validity of this analysis method has not yet been confirmedby experiments. PWRI is now conducting experimental research into the caisson
foundation of low steel reinforcement.
Areas Requiring Retrofit and Retrofit Details
Based on the seismic performance verification results described above, Fig. 17
shows a summary of the areas requiring retrofit and the details.
As for the main tower, its width-thickness ratio parameter had to be improved,
and increasing flexural strength at its tower base is necessary. For these reasons, we
adopted a strengthening method involving installing stiffing ribs bolted on the exterior
surface and these be continuous at cross ribs and diaphragms.
For the wind bearing, two methods were compared: 1) replacing the bearing,and 2) installing a structure limiting excessive displacement. Finally the method
involving the replacement of only the upper bearing was adopted from an economic
reason.
Regarding the piers, P11 and P13 incurred bending damage solely due to theinertia force due to their own weight, making it difficult to adopt retrofit methods bymaking use of dispersion of inertia force and seismic isolation. Consequently, the RC
piling method was adopted for all three piers. For its construction method, a309

Main tower :Installed ribs by the outside

Wind bearing :Replace only the upper shoeWind bearing :Replace only the upper shoe△△▽Pier :
RC piling retrofit

Pier :
RC piling retrofit

Pier :
RC piling retrofit

Blocked water by steel boxes

Blocked soil by steel pipe
sheet pile
P11

△

P12

P13

Fig.17 Seismic retrofit measurescomparative investigation was conducted between the PC confined method thatallowed underwater placement, and the method involving blocking of water by steel
casings to allow work to be performed in a dry condition. The latter method was
adopted because of the significant advantages in terms of both construction period andcost and low impact on water quality.
Conclusion
The seismic performance of the cable-stayed bridge in Sakitama Bridge was
examined. Retrofit was deemed necessary for the tower, wind bearing, and piers toensure the target seismic performance. For the caisson foundation, however, thenecessity of the retrofit will be re-examined after the results of experimental research
now conducted by PWRI becomes available.
Acknowledgments
We highly appreciate "Sakitama Bridge Seismic Retrofit Examination
Committee (Chair, Shigeki Unjou, Earthquake Disaster Researcher, NILIM)" for theirvaluable opinion and comments.
References[1] Ministry of Land, Infrastructure, Transport and Tourism: Three-Year Program ofSeismic Retrofit of Bridges on Emergency Routes, June 2005 (in Japanese).
[2] Japan Road Association: Design Specifications for Highway Bridges, Part V
Seismic Design Edition, Maruzen, March 2002.

310

REPAIR OF HIGH SHEAR STANDARD REINFORCED CONCRETE
BRIDGE COLUMNS USING CFRP
Ashkan Vosooghi1 and M. Saiid Saiidi2
Abstract
Two circular, high shear RC columns were designed identically using current
bridge seismic codes. They were damaged to near failure using one of the shake
tables at the University of Nevada, Reno. The columns were repaired using
unidirectional carbon fiber reinforced polymer (CFRP) jacketing and retested toevaluate the repair performance. The loose concrete was removed and the spalled
area was repaired using a fast-set non-shrink mortar. The cracks were epoxy injected.
Different number of CFRP layers and different repair mortar and application method
were used for the columns. The results indicate that the strength and drift capacity of
the columns were fully restored. In addition, it was found that spirals maintainapproximately 50% of their capacity even when the column damage is severe.
Introduction
Past effort in seismic design of concrete bridges has been on detailing ofbridges to prevent collapse. During earthquakes, reinforced concrete bridge columns
are designed to undergo cracking, spalling, and yielding of steel and provide
significant rotational capacity at plastic hinges so that the integrity of the overall
structure is maintained. With proper design and construction, this objective can be
met. However, the serviceability of the bridge after the earthquake is in question.
Rapid and effective repair methods are needed to enable quick opening of the bridge
to minimize impact on the community.
As part of this study, two 1/3 scale high shear standard RC bridge columns,
which were damaged to the highest repairable level in the previous tests, were
repaired using CFRP wrapping. At this level of the damage, many spirals andlongitudinal bars are visible, some of the longitudinal bars are beginning to buckle,
and the edge of concrete core is damaged. No bar is ruptured.
Column Models
NHS1 and NHS2, New-design High Shear, were the two identical doublecurvature column models that were studied. The double-curvature configuration
allows for the application of a relatively high shear, resulting in extensive shear cracks
1

PhD Candidate and Research Assistant, University of Nevada, Reno, US

2Professor, Department of Civil Engineering, University of Nevada, Reno, US

311

in concrete that induced high demand in the transverse steel, in addition to flexural
cracks and plastic hinging. Note that the columns were flexure dominated andexpected to be ductile. The latest Caltrans Seismic Design Criteria, SDC version 1.4(2006), was used to design the column. Scale factor of 1/3 was selected based on the
typical cross section dimensions of bridge columns. The scale factor was applied in a
way that stresses would not be scaled and real concrete and steel could be used.
NHS1 and NHS2 were repaired using fast set non-shrink concrete, epoxy injection,
and CFRP wrapping. Different concrete repair methods and jacket layers wereapplied for NHS1 and NHS2. The former repaired column and the latter one arecalled NHS1-R and NHS2-R, respectively. The column specifications are listed inTable 1. The test setup and column section are shown in Fig. 1. The primary test
variable was the number of CFRP wraps. The objective was to determine if byreducing the number of wraps and counting on partial contribution of the column
spiral to shear would lead to satisfactory performance.
Repair Design
The repair system was designed with the objective of restoring confinement
and shear strength of the columns by using unidirectional CFRP jacketing.
Restoring confinement
Because there are no seismic repair design guidelines available, seismic
retrofit guidelines in the California Department of Transportation (Caltrans)
provisions for RC columns were used to restore confinement using FRP jacketing.
Based on the provisions, for regions inside a plastic hinge region, without a lap splice,
it is necessary to provide a minimum confinement stress of 300 psi (2.07 MPa) at aradial dilating strain of 0.004. For regions outside the plastic hinge region, the criteria
may be reduced to a minimum confining stress of 150 psi (1.03 MPa) at a radial
dilating strain of 0.004. The length of the plastic hinge zone is defined as 1.5 times
the cross sectional dimension in the direction of bending.
The required jacket thickness is calculated as follows:

tj =

fl D
2E j ε j(1)

Where t j is jacket thickness, f l is confinement stress, D is column diameter,
E j is CFRP modulus of elasticity, and ε j is dilating strain as defined above.
Restoring shear strength

Priestley (1996) recommended that in calculating the shear resistancecontributed by the FRP, the stress in the FRP shall be limited to 0.004 E j for a strain
limit of 0.004 to avoid degradation in concrete aggregate interlock. Combining therecommendation and the Caltrans criteria for seismic shear design for ductile concrete
members, the required thickness for the jacketing, t j , is determined as:

312

Vo
tj =

φ − (Vc + Vs )(2)

π × 0.004 ×E j ×D
2

Where, Vo is over strength shear, Vc is the concrete shear capacity, Vs is theshear strength provided by spirals, and φ is 0.85. Other parameters were definedpreviously. Vo was assumed to be associated with the maximum moment achieved in
the NHS1 and NHS2 tests. Different assumptions were made for inside and outside
the plastic hinge zone to calculate Vc and Vs :Inside the plastic hinge
Since some of the thin cracks are not repairable inside the core, Vc wasneglected in the both columns. The spirals for NHS1 experienced a strain greater than
1.75 yield strain. As a result, Vs was assumed to be zero for NHS1-R. Testing
NHS1-R on a shake table and calculating the contribution of the spirals to shear
showed that spirals resisted approximately 50% of the shear even though they were
neglected in design. In NHS2-R this led to a smaller number of CFRP Layers.

Outside the plastic hinge
Since the spirals do not yield outside the plastic hinge, a 100% credit wasgiven to spirals in both columns. Although shear cracks occurred outside the plastic
hinge too, the level of damage was much lower than that of the plastic hinge. As aresult, 50% credit was also given to the concrete outside the plastic hinges. A jacket
system with a thickness of 0.04 in (1.0 mm) per layer was used.

Repair Process
Neglecting the unexpected delays, the entire repair process can be conductedin one day. The repair process is shown in Fig. 2 for one of the plastic hinges. The
repair process consisted of the following steps:

Straightening the column
By adjusting the shake table, the column was returned to the initial vertical position.

Removing loose concrete
The loose concrete was removed by an impact hammer with a chisel point. Mostly,
the loose concrete was removed from the compression dominant side of the column.

Concrete repair
Two different types of mortars and mortar placement were used for NHS1-R
and NHS2-R. In NHS1-R, a one component, micro silica and latex modified, and nonsag repair mortar was used. The specified 3-day compressive strength of the mortar in
NHS1-R was 4000 psi (27.6 MPa). In NHSR-2, a rapid repair mortar was used. The

313

mortar was low-shrinkage, microsilica-modified, cement-based mortar for structural
repair or overlays. The specified 3-hour and 1-day compressive strengths of themortar used for NHS2-R were 3000 psi (20.7 MPa) and 4000 psi (27.6 MPa),
respectively.

Pressurized epoxy injection of the cracks
The epoxy was injected into shear cracks and flexural cracks on the tension
dominant side of the column. To inject the epoxy, an inlet was put at one end of acrack and an outlet was put at the other end. Then the surface of the crack wascovered by removable glue. Epoxy was injected with a standard pressure of 40 psi(0.28 MPa) from the inlet until it bled from the outlet to ensure that the crack wascompletely filled with epoxy.

Surface preparation for CFRP wrapping
Column surface was roughened slightly by a grinder. A layer of epoxy wasapplied to prime the columns surfaces. After that, a thickened epoxy was applied
directly to the columns to smooth out imperfections.

CFRP wrapping
After preparing the surface, the epoxy was applied to CFRP layers using apaint roller. CFRP layers were wrapped around the columns manually.

Curing
The entire curing took less than 48 hours which was composed of first half ofaccelerated curing, followed by curing under the lab ambient condition. During
accelerated curing, the temperature was elevated to 100 o F to 112 o F , and the relative
humidity was reduced to 10% to 15% .

Test Protocol

The Sylmar ground motion record was applied in shake table testing of thecolumns. The record was applied to the columns with amplitudes increasing
gradually. The objective of NHS-1 and NHS-2 testing was to reach to the imminent
failure state. In this damage state, the column is approaching failure and damage has
begun to penetrate the confined core. No bar rupture is desired in this damage state.
To evaluate the repair performance, the repaired columns were subjected to identicalincreasing motions as original columns with additional motions having higher
amplitudes applied to the repaired columns until failure.
Shake Table Tests Results

The columns were tested on one of the shake tables for University of Nevada,Reno. The failure mode for NHS1-R was fracture of two longitudinal bars at the base(Fig. 3). In NHS2-R failure was due to the CFRP rupture at the column base on thecompression dominant side along with rupture in two longitudinal bars at the base(Fig. 4). In both columns, bar ruptures were noted by removing the CFRP jacket.

314

The cumulative force-displacement hysteresis and back bone curves for
NHS1-R and NHS2-R are shown in the upper graphs in Fig. 5. In the lower left
graph, back bone curves for NHS1 and NHS1-R are plotted. Those of NHS2 and
NHS2-R are plotted in the lower right graph.
Strength, stiffness, and deformability are the main characteristics of astructure. To compare the performance of the original columns and repaired ones, thefallowing non-dimensional response indices were developed. These indices can also
be used among repaired columns to compare the performance of different repair
methods.
Strength Index (STRI)

The lateral strength of a column is defined as the peak measured base shear.
The ratio between the lateral strength of the repaired column and the original one isdefined as strength index:
STRI =

VR
VO(3)

Where, VR and VO are the peak base shears for the repaired column and theoriginal one, respectively.
Stiffness Index (STFI)The serviceability of a repaired structure also needs to be considered. The
stiffness of the structure under low amplitude lateral loads is an important parameterfor quantification the serviceability. Assuming a point on the push over curve with
one-half of the peak base shear as the elastic limit, the chord service stiffness of thecolumn is calculated by dividing the one-half of the peak base shear by thecorresponding displacement. Having the chord service stiffness for the originalcolumn and the repaired one, the stiffness index is found as follow:
STFI =KR
KO

Where, KR and KO are the chord service stiffness for the repaired column andthe original one, respectively. It should be noted that the elastic limit of the repaired
column is not taken larger than that of the original column.
Deformability Index (DI)This index is defined as the ratio between the drift capacity of the repaired
column (DR ) and that of the original column (DO). Deformability index is determined
as follows:
DI =

DR
DO

(4)

315

Since the original columns were not tested to failure, their drift capacity islarger than the maximum measured drift. Past failure test data have shown that forwell designed columns under high shear, the ratio of ultimate displacement todisplacement at imminent failure is approximately 1.2. As a result, to calculate DI,
the maximum measured drifts for NHS1 and NHS2 were increased by 20%.
The peak base shear, maximum drift, and service stiffness for the columns arelisted in Table 2. These parameters were used to calculate the response indices.
Although the strength of NHS2 was less than that of NHS1, the repaired columns had
equal strengths. NHS2-R had considerably higher service stiffness than that of
NHS1-R, however those of NHS1 and NHS2 were almost the same. The table also
shows that the maximum drift for both repaired column was nearly the same.
The response indices are plotted in Fig. 6. In general, all the response
indices for NHS2-R are higher than those of NHS1-R. It implies that the repair
procedure in terms of quality and application method of the repair mortar has
significant role in the performance of the repaired column. In addition, the number of
CFRP layers in NHS1-R was not optimized and giving 50% credit to the spirals for
NHS2-R was a reasonable assumption. Fig. 6 shows that the strength anddeformability of both columns were fully restored. Although the service stiffness was
not fully restored in both columns, but the stiffness reduction in NHS2-R was 2/3 ofthat of NHS1-R. The reason is that the higher quality repair mortar and better
application method, pouring and vibrating instead of patching, were used in NHS2-R.
Conclusions

Based on the observations and the measured data from the testing of theoriginal and the repaired columns, the following conclusions are made:-The repair design method was rapid, and effective because it restored the lateralload and drift capacity of the columns.

-

The repair process was practical and may be used for rapid emergency repair ofearthquake damaged concrete columns.

-

Giving 50% credit to the spirals capacity and neglecting the concrete strength
inside the plastic hinges is a reasonable assumption in the repair design.

-

Giving full credit to the spirals capacity and 50% credit to the concrete strength
outside the plastic hinges is a reasonable assumption in the repair design.

References
1- Ang Beng Ghee, M. J. N. Priestley, and T. Paulay, “Seismic Shear Strength of
Circular Reinforced Concrete Columns,” ACI Structural Journal, January-February
1989, pp. 45-59.

316

2- California Department of Transportation, “Memo to Designers 20-4 attachment
B,” California, USA: Engineering service center, earthquake engineering branch,
2007.
3- California Department of Transportation, “Seismic Design Criteria, version
1.4,” California, USA: Engineering service center, earthquake engineering branch,
2006.
4- California Department of Transportation, “Bridge Design Specifications,”
California, USA: Engineering service center, earthquake engineering branch, 2003.
5- Johnson, N., M. Saiidi, A. Itani, and S. Ladkany, “Seismic Retrofit of Octagonal
Columns with Pedestal and One-Way Hinge at Base,” ACI Structural Journal, Vol.
102, No. 5, 2005, pp. 699-708.
6- Laplace, P. N., D. H. Sanders, M. Saiidi, B. M. Douglas, and S. El-Azazy,“Retrofitted Concrete Bridge Columns Under Shake Table Excitation,” ACI
Structural Journal, Vol. 102, No. 4, 2005, pp. 622-628.
7- Li, Y. F. and Y. Y. Sung, “Seismic Repair and Rehabilitation of a Shear-Failure
Damaged Circular Bridge Column Using Carbon Fiber Reinforced Plastic Jacketing,”
Canadian Journal of Civil Engineering, No. 30, 2003, pp. 819-829.
8- Priestley, M. J. N., F. Seible, and G. M. Calvi, “Seismic Design and Retrofit ofBridges,” New York, USA: John Wiley & Sons, 1996.
9- Saiidi, M. and Z. Cheng, “Effectiveness of Composites in Earthquake Damage
Repair of Reinforced Concrete Flared Columns,” Journal of Composites forConstruction © ASCE Vol. 8, No. 4, 2004, pp. 306-314.
10- Saiidi, M., K. Sureshkumar, and C. Pulido, “A Simple Model for CFRP
Confined Concrete,” Journal of Composites for Construction, ASCE, Vol. 9, No. 1,
2005, pp. 101-104.
11- Teng, J. G., J. F. Chen, S. T. Smith, and L. Lam, “FRP Strengthened RC
Structures,” Sussex, England: John Wiley & Sons, Ltd., 2002.
12- Vosooghi, A., M. Saiidi, and J. Guitierrez, “Rapid Repair of RC Bridge
Columns Subjected to Earthquakes,” 2nd International Conference on Concrete
Repair, Rehabilitation, and Retrofitting, Cape Town, S. Africa, November 2008, pp.
1113-1119.

317

Table 1. Specifications for NHS1 and NHS2
H
in (mm)Din (mm)80 (2032)

16 (406)

Long.
Steel Ratio

Trans.
Steel Ratio

3.08%

1.34%

Aspect
Ratio

Axial Load,kips (kN)

2.5

100 (444.8)

Table 2. Main responses of the columns

NHS1
NHS1-R
NHS2
NHS2-R

Peak base shear,
Kips (kN)
94.1 (418.6)
95.2 (423.6)
78.9 (350.9)
92.1 (409.7)

Maximum drift
9% [1.2*7.5%]
13.1%
7.7% [1.2*6.4%]
13.3%

Service stiffness,
Kips/in (kN/mm)
73.2 (12.8)
28.3 (5.0)
71.0 (12.4)
44.0 (7.70)

Fig 1. Test setup and section propertis for NHS1 and NHS2318

2a) Before repair2b) Concrete Chipping2c) Concrete patching (NHS1-R)

2d) Smoothening (NHS-R)2e) Conc. pouring, vibrating (NHS-R)

2g) Epoxy injection2h) Surface preparing

2f) After concrete repair2i) CFRP wrappingFig 2. Repair process for NHS1 and NHS2

Fig 3. Bar ruptures in NHS1-R. (CFRP was removed after shake table tests.)

319

(a)

(b)

Fig 4. Failure in NHS2-R; a) CFRP rupture; b) bar rupture (CFRP was removed after
shake table tests.)
Displacement [mm]152

NHS1-R

-2

0

2

4

6

Displacement [mm]

203

254

305
534
445
356
267
178
89
0
-89
Back Bone -178
Measured
-267
-356
8
10
12

-51
120
100
80
60
40
20
0
-20
-40
-60

0

-2

0

203

120

305
534

100

445

80

356

60

267

40

178

NHS1-R
NHS1

20

89

0

0

2

4

6

8

4

6

10

51

102

152

305
534

100

445

80

356

60

267
178

12

2

4

6

1.46

1.17
1.01

1
0.62

0.6

0.39

0.4
0.2
0
STRI

STFI

8

Displacement [in]

1.73

0.8

89
0

0

1.4
1.2

254

NHS2-R
NHS2

20

2
1.6

203

40

Fig 5. Lateral force-displacement relationships for the columns
NHS1-RNHS2-R

305
534
445
356
267
178
89
0
Back Bone -89-178
Measured
-267
8
10
12

120

Displacement [in]

1.8

254

0

0
0

2

203

Displacement [mm]254

Lateral Force [kips]

152

Lateral Force [kN]

Lateral Force [kips]

102

152

Displacement [in]Displacement [mm]51102

NHS2-R

Displacement [in]051Lateral Force [kN]

102

DI

Fig 6. Response indices for the repaired columns

320

10

12

Lateral Force [kN]51

Lateral Force [kips]

120
100
80
60
40
20
0
-20
-40
-60
-80

0

Lateral Force [kN]

Lateral Force [kips]-51

SEISMIC RETROFIT TECHNIQUES FOR REINFORCED CONCRETE BRIDGECOLUMNS WITH COMBINATION OF FRP SHEET AND STEEL JACKETING
Guangfeng Zhang1, Shigeki Unjoh2, Jun-ichi Hoshikuma3 and Junichi Sakai4
Abstract
With the purpose to provide a seismic retrofit method with good workability forreinforced concrete (RC) highway bridge columns under severe construction work
conditions, a retrofit method with combination of carbon fiber reinforced polymer (CFRP)
sheet and steel jacketing has been proposed in the previous researches of this research
project. In this research, bonding behavior between CFRP sheet and steel plate wasinvestigated experimentally in order to establish a rational design method for the
CFRP-steel bonded connection. This paper provide an introduction of the proposed
seismic retrofit method companying with descriptions of the tests on CFRP-steel bonded
connection and test on RC bridge column specimen for examining the retrofit effectivenessof the proposed retrofit method.
Introduction
Attributing to the merits of high strength, light-weight and outstanding workability,
fiber reinforced polymer (FRP) sheet has been widely used in repairing or strengthening
reinforced concrete (RC) members in the recent decades. As for seismic retrofit of RC
highway bridge columns in Japan, FRP sheet is usually used to retrofit columns withpremature termination of longitudinal reinforcements without enough development length
at the midheight. Figure 1 shows an example of RC bridge being under retrofitted with
Carbon FRP (CFRP) sheet. CFRP sheet was jacketed around the termination sections of
the longitudinal reinforcements in the longitudinal direction and circumferential direction
to reinforce flexural and shear strength. In some other cases, FRP sheet is also used toreinforce ductility of the columns by jacketing around the plastic hinge in thecircumferential direction.
However, it is not an effective method to reinforce the flexural strength of thecolumn base by jacketing FRP sheet in the longitudinal direction. Generally, highelongation is required locally in both of the longitudinal and circumferential directions at
the base. Enhancement of ductility capacity of the base can not be obtained because
elongation of the FRP sheet is rather lower before breaking. From the point of view, a
1Researcher, Bridge and Structural Technology Research Group, Public Works Research Institute, Japan
Research Coordinator for Earthquake Disaster Prevention, Research Center for Disaster Risk Management,
National Institute for Land and Infrastructure Management, Japan
3
Chief Researcher, Bridge and Structural Technology Research Group, Public Works Research Institute,Japan
4Senior Researcher, Ditto
2321method with using combination of CFRP sheet and steel jacketing has been proposed in theprevious researches (Zhang and Unjoh, 2009). It should be noted for the retrofit methodthat bonding behavior between CFRP sheet and steel plate is a key issue, because thelongitudinal force induced from the additional anchor bolts must be transmitted to CFRP
sheet and thus the steel plate should be bonded with CFRP sheet in the inelastic responseof the column.
This paper provide an introduction of the proposed seismic retrofit method
companying with descriptions of the tests on CFRP-steel bonded connection and test on
RC bridge column specimen for examining the retrofit effectiveness of the proposed
retrofit method.

Fig. 1 Retrofit of RC bridge columns with CFRP sheet

Overview of Seismic Retrofit methods
Figure 2(a) shows schematic of the proposed retrofit method. It adopted a
combination of CFRP sheet jacketing and steel plate jacketing (hereinafter, combination of
CFRP and steel jacketing) taking advantage of the merits of both of the CFRP and steel.
Considerations of this technique are to use different methods to retrofit the base and theother parts of a column. A method same as the normal steel jacketing is used to retrofit thebase where upgrading of the strength and the ductility are required. A FRP jacketing isused to retrofit the other parts to provide high strength. Transmit of force between the steel
plate and the CFRP sheet is accomplished by FRP-steel bonded connection with a certain
bond length.
Comparing to the normal steel jacketing and RC jacketing as shown in figure 2(b)322and (c), the proposed method is complicated because it needs to design the details of thebonded connection. The merits of the proposed method are the high performance ofconstruction and that increasing of the death weight after retrofit is very limited. The
proposed method be suitable for the situations such as the followings: 1) bridges located ina place where no enough construction space can be ensured for building a large scale
scaffolding or for the working of the construction machines and equipments; 2) bridges
crossing a river or bridges located in a heavy snows region where construction period isshort, etc.
Because this method employs a FRP-steel bonded connection to transfer thestrength upgraded by the anchor bolts to the CFRP sheet, it is important to ensure thebonding capacity and certainty of the bonded connection. In this research, bonding
behavior between CFRP sheet and steel plate was investigated intensively with shear bondtests in order to establish a rational design method for the bonded connection. Furthermore,a RC bridge column specimen was tested under cyclic loading method to inspect theretrofit effectiveness of the proposed retrofit method.

(a) Combination of CFRP
and steel jacketing(b) Steel jacketing(c) RC jacketing

Fig. 2 Schematics of seismic retrofit methods for RC bridge columns

Tests on Bonding Behavior between CFRP Sheet and Steel Plate
Four series of double-lap shear bond tests were carried out taking the bonded layersof CFRP sheet as parameter. Three specimens were tested for each series. Table 1 showsa list of the specimens.
Figure 3 shows the details of the double-lap shear specimens. CFRP sheet of 600
g/m2 and ss400 steel plates were used. A long bond length of 600 mm was set in the tests
in order to inspect the progress of debonding between CFRP sheet and steel plate. Surfaceof steel plate to be bonded with CFRP sheet was grinded with disc sanding machine. All323

the specimens were cured for 7 days at ambient temperature. The tests were performed
using a universal testing machine with a loading speed of 0.5-1.0 mm per minute.
Table 2 shows the test results of material properties of CFRP sheet and epoxy
adhesive.

Series
1-layer2-layer
3-layer4-layer

Materials
CFRP sheet

Epoxy adhesive

Table 1 List of Specimens
Bond length
Bonded layers of
(mm)
CFRP
600
1
600
2
600
3
600
4

Table 2 Material Properties
Items
Mass per area (g/m2)
Design thickness (mm)Young’s modulus (N/mm2)
Tensile strength (N/mm2)
Young’s modulus (N/mm2)
Compressive strength (N/mm2)
Tensile strength (N/mm2)
Tensile shear strength (N/mm2)

Numbers of specimens
for each series
3 (No.1, No.2, No.3)
3 (No.1, No.2, No.3)
3 (No.1, No.2, No.3)
3 (No.1, No.2, No.3)

Values600
0.333
2.60×105
3,400
2.202×103
71.8
47.0
17.6

Fig. 3 Details of specimens and setup of test

324

Fig. 4 Examples of test results on bonding behavior between CFRP sheet and steel plate
Test results were summarized with the items of relationship between shear force
and displacement of CFRP at the end of steel plate, distribution of strain in CFRP andinterfacial bond stress along the bond length. Here, bond stress was calculated from thestrain using the relationship τ = dε ( x) dx ⋅ E f ⋅ t f , where, E f and t f are the Young’s
modulus and thickness of CFRP sheet. Figure 4 shows test results of a 1-layer specimen
and a 4-layer specimen for examples.
The value of strain at the point of debonding initiated and the length of effective
bond length were discussed. The value of strain at the point of debonding initiated will
become lower with increasing of the bonded layers. The lowest value was confirmed as
about 3,500 microstrain for 3-layer specimen and 4-layer specimen. The length of effective
bond length is about 80-100mm for 1-layer specimens, 100-160mm for 2-layer specimens,
120-180mm for 3-layer specimens and 140-200mm for 4-layer specimens. From the viewof design, it can be said from these results that bond length and strain of CFRP sheet usedin design should be considered carefully to ensure the design strength.

325

Fig. 5 Details of RC bridge column specimen before and after retrofit withcombination of CFRP and steel jacketing

Cyclic Loading Test on RC Bridge Column Specimen
A RC bridge column specimen was tested under cyclic loading method to inspect
the retrofit effectiveness of the proposed combination of CFRP and steel jacketing method.
The specimen was a scaled model of a RC bridge column with premature termination of
longitudinal reinforcements at the midheight but retrofit of the termination section of thelongitudinal reinforcements was already performed. Retrofit of the termination section of
the longitudinal reinforcements was designed to prevent failure at this section period toflexural failure of the base. Therefore, it can be considered that the specimen before
retrofitted with the proposed method is equivalent to a column with a failure mode offlexural failure of the base.
Figure 5 shows the details of the specimen. Cross section of the original column
was 600×600 mm. The height of the loading point was 3,100 mm with an aspect ratio of 5.0.
SD295-D10 and SD295-D3 was used as longitudinal reinforcements and hoop ties,
respectively. Reinforcement ratio was 1.58%. Retrofit with the proposed retrofit method
was designed to upgrade the flexural strength with an increase of 30 percent. Bond lengthof 350 mm was applied between CFRP sheet and steel plate.
Figure 6 shows setup of the cyclic loading test. Axial force of 539 kN was applied
during the loading to provide an axial stress of about 1.5MPa. Loading was performed
under displacement control and the loading speed was 10 mm/sec. Cyclic number was 3 foreach loading step.

326

Fig. 6 Setup of cyclic loading test

Fig. 7 Conditions of specimen after test
Cyclic loading was surcharged until a loading step of 9 δ y with yielddisplacement δ y of 12.7 mm. Initiation of cracks at the base, breaking of anchor bolts,
remarkable compressive failure of core concrete and breaking of longitudinal
reinforcements were confirmed at a loading step of 3 δ y , 6 δ y , 7 δ y and 8 δ y , respectively.
Bonding condition between CFRP sheet and steel plate was checked after each loading
step by tapping the CFRP surface with a small hammer to detect voids. It is noted that nodebonding area was detected till the test finished. Figure 7 shows the conditions ofspecimen after test.

327

Fig. 8 Lateral force-displacement
Retrofit effectiveness of the proposed retrofit method was discussed by comparing
the flexural strength and ductility with that of a specimen retrofitted with normal steel
jacketing and that of a reference specimen (Hoshikuma, J. and Unjoh, S.). Here, design ofthe base section of the original column was the same for each of the three specimens. And
design of the anchorage was the same for each of the two retrofitted specimen. Figure 8
shows lateral force-displacement hystereses and envelope curves of the three specimens. It
can be seen that retrofit effect of the proposed retrofit method was similar with the normal
steel jacketing.
Conclusions
Retrofit method with using a combination of CFRP sheet jacketing and steel
jacketing was proposed in this research project with the purpose to provide a seismic
retrofit method for RC bridge columns under severe construction work conditions.
Bonding behavior between the CFRP sheet and steel plate was investigated experimentallyin order to establish a rational design method for the bonded connection. It is found that the
proposed retrofit method with necessary bonding connection details can provide a retrofit
effect similar with the normal steel jacketing.

328

References
Zhang, G.F. and Unjoh, S.: “A Retrofit Method for Upgrading Seismic Performance of RC
Columns with Termination of Main Reinforcements”, Proceedings of the 12th Symposiumon Ductility Design Method for Bridges, JSCE, pp.47-52, 2009.1 (in Japanese)
Hoshikuma, J. and Unjoh, S.: “Seismic Retrofit of Existing Reinforced Concrete Columns
by Steel jacketing”, Proceedings of the Second Italy-Japan Workshop on Seismic Design
and Retrofit of Bridges, Rome, Italy, February 27-28, 1997, pp.413-428.

329

blankDEVELOPMENT AND REFINEMENT OF ILLINOIS’
EARTHQUAKE RESISTING SYSTEM STRATEGY
Daniel H. Tobias1, Jerome F. Hajjar2, Ralph E. Anderson1, James M. LaFave2, Chad
E. Hodel3, Larry A. Fahnestock2, William M. Kramer1, Joshua S. Steelman2, Patrik D.
Claussen1, Kevin L. Riechers1, and Mark D. Shaffer1
Abstract
An important aspect of newly adopted American Association of State
Highway and Transportation Official’s (AASHTO) bridge design code provisions is a
design earthquake with significantly increased accelerations. The Illinois Departmentof Transportation (IDOT) has recently developed and implemented an Earthquake
Resisting System (ERS) strategy for all bridges in the state in order to accommodate
the increased AASHTO seismic design hazard. This paper provides a short history ofthe development of IDOT’s ERS strategy as well as an overview of a recently
established research program aimed at refinement and calibration at the University of
Illinois at Urbana-Champaign (UIUC).
Introduction
BackgroundIn 2008 and 2009, the American Association of State Highway andTransportation Officials (AASHTO) published modernized codified standards for thedesign of highway bridges to resist earthquake loadings. The revised and updated
provisions are contained in the AASHTO Load and Resistance Factor Design Bridge
Design Specifications (LRFD Code) (AASHTO 2009b) and the 1st edition of the
AASHTO Guide Specifications for LRFD Seismic Bridge Design (LRFD Seismic
Guide Specifications) (AASHTO 2009a). These two documents reflect theculmination of several years of effort by the bridge engineering community in the
United States (MCEER/ATC 2003; NCHRP 2006). Both the LRFD Code and the
LRFD Seismic Guide Specifications have incorporated a design earthquake with a1000 yr. return period (Leyendecker et al. 2007). Prior to 2008, the codified design
return period earthquake was 500 yrs. (FEMA 1988; AASHTO 2002). The methods
and soil parameters used to determine design earthquake response spectra (BSSC
1995) along with numerous other aspects of seismic bridge design philosophy werealso modernized in the recently published AASHTO documents.
Traditionally, the philosophies and need to address typical bridge
configurations in western states within the United States have driven advancements
and codified provisions for seismic design of highway bridges nationwide. Seismic
considerations have been a primary concern in these states for many years due to the1

Illinois Department of Transportation, Springfield, Illinois, USA.
University of Illinois at Urbana-Champaign, Urbana, Illinois, USA.
3
WHKS & Company, Springfield, Illinois, USA.
2

331

region’s widely recognized high risk for large damaging earthquakes. With theadvent of the newly adopted 1000 yr. design return period earthquake, though, there isa keener recognition of the risk for a large seismic event happening in some midwestern and eastern states. Significant earthquakes in these regions of the UnitedStates are known to have occurred, but the frequency of recurrence can be quite long.
In the last several years, many states east of the Rocky Mountains, including Illinois,
have made significant strides with regards to seismic design, retrofitting andconstruction methods for highway bridges. The modernization efforts in Illinois havefocused on seismic design philosophies and methods that are the most appropriate fortypical bridge configurations constructed in the state.
Scope of Increased Design Hazard
Designing bridges for a 1000 yr. earthquake, which primarily affects
approximately the southern half of Illinois, represents a significant increase in design
accelerations from the 500 yr. event. Fig. 1 presents an approximate 1000 yr. spectral
acceleration map for Illinois at a period of 1.0 sec. assuming all the soil is classified as
Soil Site Class D according to the definitions contained in the LRFD Code and the
LRFD Seismic Guide Specifications. Site Class D is a common soil type in southern
regions of Illinois. The figure gives a generalized idea of the seismicity of Illinois forthe 1000 yr. design seismic event. Spectral accelerations at 1.0 sec. are used todelineate between Seismic Performance Zones (SPZ) in the LRFD Code and Seismic
Design Categories (SDC) in the LRFD Seismic Guide Specifications. As the SPZ(and SDC) increases in number (letter), so do the seismic design requirements. For
the 500 yr. design earthquake, there were no parts of Illinois in SPZ 4 and only asmall portion was in SPZ 3. Furthermore, the design accelerations within each SPZ
are significantly greater for the 1000 yr. earthquake as compared to the 500 yr.
earthquake.
Development and Refinement
In late 2006 and mid 2008, after several years of development, the Illinois
Department of Transportation (IDOT) published and implemented initial versions of acomprehensive strategy or framework for the design, retrofit and construction ofbridges to resist seismic loadings in the IDOT Bridge Manual (IDOT 2008a) and the
IDOT Seismic Design Guide (IDOT 2008b). The strategy is comprised of a set ofcore concepts and structural details which, when implemented together, form ageneralized Earthquake Resisting System (ERS). Illinois’ ERS strategy is flexible
enough to be applicable to all common bridge types built in the state, and for any pastor future codified hazard level. Pertinent aspects of recently adopted seismic
provisions in the LRFD Code, the LRFD Seismic Guide Specifications, the Federal
Highway Administration Seismic Retrofitting Manual for Highway Bridges (FHWA
Retrofit Manual) (FHWA 2006), and several other sources (ICC 2000; AASHTO
2000) were used to formulate and tailor a viable ERS framework for the state. Rfactor or forced based concepts (from the LRFD Code) are used as a primary basis fordesign.

332

FIG. 1. APPROXIMATE SPECTRAL ACCELERATIONS AT A PERIOD OF 1.0
SEC. FOR SITE CLASS D SOIL IN ILLINOIS
Initial modernization and improvement efforts by IDOT in the state coincidedwith those of AASHTO over the last several years (Hodel et al. 2004; Tobias et al.
2006a; Tobias et al. 2006b; Tobias, et al. 2008a; Tobias, et al. 2008b). The processesof updating the LRFD Code and formulating the LRFD Seismic Guide Specifications
were undertakings in which Illinois played an important role as a contributing
member of the AASHTO Technical Subcommittee on Seismic Bridge Design (T-3).
The involvement by IDOT at the national level helped to greatly enhance the locallydeveloped Illinois ERS strategy. IDOT had the opportunity to garner experience andexpertise from other contributing states, and also played a role in crafting some of the333key provisions that pertain to states with typical bridge configurations and seismicities
which are similar to Illinois.
A research program initiated in early 2009 by IDOT at the University of
Illinois at Urbana-Champaign is primarily focused on refinement and calibration of
the currently implemented forced based approach for areas in Illinois with moderate
to moderately high seismicity. In time, for areas in Illinois with higher seismicity,
additional aspects of the recently developed displacement based approach (from the
LRFD Seismic Guide Specifications) are also anticipated to be incorporated into
Illinois’ ERS strategy through the results of the ongoing research effort.
The first part of this paper provides a brief overview of the initial developmentof Illinois’ ERS strategy over the last 5 to 6 years, while the second part provides a
description of the ongoing research program aimed at calibration and refinement.
ERS Strategy
Range of Applicability
Illinois’ ERS strategy is primarily intended for common bridge types built inthe state that normally can be designed assuming the first mode of vibration is thedominant response to a seismic loading (i.e., are regular as defined by the LRFD Codeand the LRFD Seismic Guide Specifications). Superstructures of these bridge
configurations generally include those with a concrete deck on steel or precast
prestressed I-beams, or a wearing surface on precast prestressed deck beams (boxbeams). The abutments include non-integral and integral stub. The piers can be ofvarious types including multiple column concrete bents, several variations of drilledshaft bents, and solid walls. Foundation types include spread footings, HP and metalshell piling, and drilled shafts. Various combinations of these elements make up thevast majority of Illinois’ inventory at present and for the foreseeable future. For
bridges that are irregular, the general principles of the Illinois ERS strategy are also
applicable, and required to be followed. Irregular bridges typically require multimodal design and analysis methods.
Seismic Structural Redundancy Levels
The underlying philosophy of Illinois’ ERS strategy is to allow certain levelsof damage during a seismic event at planned locations in a structure such that loss of
span is prevented. Loss of span directly impacts critical public transportation
facilities, and can potentially lead to loss of life. Optimally, prevention of span loss isachieved through what can be termed “levels or tiers of seismic structural
redundancy” that dissipate energy from an earthquake in key components of bridges
in succession as they fail (fuse) or engage, and alter the response of a structure. These
key bridge components include weak or fuse-like connections between
superstructures and substructures of bridges, conservative beam seat widths (support
lengths) on substructures, an allowance for plastic deformation in superstructurecomponents such as steel diaphragms, an allowance for plastic embankment334

deformation at abutments, and an allowance for plastic hinging in selected parts ofsubstructures and foundations (when necessary).
The first tier or level of seismic structural redundancy, and theoreticallyweakest fuse in Illinois’ ERS strategy, is the connections between superstructures andsubstructures. These connections are designed to fail at a nominal level of dynamic
excitation while still meeting the strength requirements for normal or service loads(non-extreme event). Fig. 2 presents a typical fusible elastomeric bearing andsuperstructure-to-substructure connection with side retainers for steel I-beams used in
Illinois, and Fig. 3 illustrates the details of the side retainer design. Most non-integral
connections between superstructures and substructures for typical bridges in Illinois
are designed to nominally carry 20% of the tributary dead weight of a superstructurein the restrained direction regardless of the Seismic Performance Zone of thestructure. In 2008, AASHTO updated and clarified the provisions for the seismic
design of connections between superstructures and substructures. The notion that, atthe discretion of owner, bearings and their connections may be designed as sacrificial
elements (i.e. fusible) is now fully endorsed alongside the historical concept which
required connections between superstructures and substructures always be designed tostay elastic during a design seismic event.
For Illinois, it is much more economical and logical to embrace the fuseconcept between superstructures and substructures in order to adapt to increased
design accelerations. The benefits of energy dissipation and an increased chance of
substructure/foundation survival (possibly beyond the 1000 yr. seismic design event)outweigh the cost of modifications to typical bridge configurations in Illinois thatwould be required to keep these connections elastic during a large seismic event.
Once superstructure-to-substructure connections have fused during anearthquake, adequate seat widths for beams on substructures are provided such thatsuperstructures can “ride out” the remainder of a design seismic event (or possibly
greater). Conservatively designed seat widths are the second tier of seismic structural
redundancy in Illinois’ ERS strategy. The empirical relationship for determining
required support length, N (m), used by Illinois is given by Eq. 1.

N

0.10 0.0017 L 0.007 H

0.05 H 1

B2
L2

1 1.25 Fv S1
cos

(1)

where L = typically length between expansion bearings (m); H = tallest pier between
expansion bearings (m); B = out-to-out width of superstructure (m); B/L = ratio not tobe taken greater than ⅜; = skew angle (°); and FvS1 = one second period spectral
response coefficient modified for Site Class. The required seat widths calculatedusing Eq. 1 are typically about 25 to 30% greater than that required by the LRFD
Code and the LRFD Seismic Guide Specifications.

335

FIG. 2. TYPICAL FUSIBLE ELASTOMERIC BEARING AND CONNECTION
DETAIL WITH SIDE RETAINERS FOR STEEL I-BEAMS

FIG. 3. TYPICAL FUSIBLE SIDE RETAINER DESIGN FOR STEEL I-BEAMS
The third tier of seismic redundancy in Illinois’ ERS strategy generally
encompasses plastic hinging of elements in substructures such as reinforced concrete
columns and, when necessary due to a bridge’s configuration,
substructure/foundational elements such as piles and drilled shafts. Figs. 4 and 5
provide some typical seismic details used in Illinois for multiple round columns piersand drilled shafts. Embankments at abutments are generally considered sacrificial
elements as part of the third tier of seismic structural redundancy. The “amperagelevel” for fuses in substructures and foundations is generally somewhat greater thanthose of the first tier of seismic redundancy. Failure of the connections between
superstructures and substructures along with plastic deformations in superstructure336

diaphragm elements helps to provide an enhanced probability that the third tier of
redundancy may not become fully engaged or fuse during a moderate to significant
seismic event (but probably will during a major seismic event without causing spanloss). The concept is analogous to comparing the primary electrical fuse for an entire
house to the many secondary individual fuses that are connected to it.

FIG. 4. SEISMIC CONFINEMENT DETAILING OPTIONS FOR CIRCULAR
COLUMNSand R-factors
Depending on the specific situation, varying degrees of isolation between
superstructures and substructures are likely provided after fusing occurs primarily
because friction may be the only mechanism of seismic force transfer at theseinterfaces. If elastomeric bearings are employed on a structure, some isolation is also
provided before fusing. Since there are several sources of seismic energy attenuationin the load path from superstructures down to their interfaces with substructures,
IDOT permits an increase in some or strength reduction factors for the 1000 yr.
design return period seismic event from those prescribed by the LRFD Code. These
increases primarily apply to reinforced concrete pier construction (usually from 0.9 to1.0) for combined moment and axial force resistance. R-factors are primarily used in
the design of substructures/foundations to promote ductile structural response during
an earthquake by reducing design moments. The recommended R-factors in Illinois’337

ERS strategy for substructure and foundation design generally follow the guidanceand bounds of the LRFD Code. However, interpretation and judgment was required
to develop and clarify for practitioners values that should be used for design which areapplicable to specific pier and abutment types built in Illinois. Refinement andcalibration of the and R-factors currently used in Illinois’ ERS strategy are one ofthe primary focuses of the research effort that is currently ongoing at the University of
Illinois at Urbana-Champaign.

FIG. 5. SEISMIC DETAILS FOR INDIVIDUAL COLUMN DRILLED SHAFT
PIERS

338

Calibration and Refinement of ERS Strategy
GeneralIDOT has teamed with the University of Illinois at Urbana-Champaign toobtain assistance with furthering the enhancement of Illinois’ ERS strategy. There aresome theoretical methods and assumptions embedded in the strategy that lack fullverification as well as significant areas which can be targeted for refinement andcalibration. An experimental university testing program that also includes theemployment of sophisticated analytical computer models can provide each of these to
IDOT. In recognition of the variability and uncertainty inherent in seismic design,
Illinois’ ERS strategy was initially formulated as a fairly conservative framework of
relatively straightforward design and detailing methods primarily intended toeconomically streamline seismic bridge design efforts. The changes to the AASHTO
bridge design provisions (primarily the adoption of the 1000 yr. design return period
earthquake) have substantially increased the population of structures in Illinois
requiring seismic analysis and design. In order to fully realize the economic benefitsof Illinois’ recently established ERS strategy, it is in the state’s best interest to make
improvements such that it will be less costly and time consuming to obtain a similar
and heightened level of seismic resistance for new bridges. In the long term, it will
also be much more economical for Illinois to achieve some level of uniformity inseismic resistance for its full population of bridges (new and retrofitted).
University of Illinois researchers are employing the current documentation of
Illinois’ ERS strategy, an oversight panel comprised of IDOT and Federal Highway
Administration engineers, information from prior research on bridge behavior in MidAmerica, as well as any other ongoing national and international initiatives related toseismic analysis and design to provide key background for the project. It is expectedthat the research effort will lead to a more rational and consistent bridge design
approach that balances structural safety with design methodologies, construction
practices, and construction costs appropriate for Illinois. In addition to more realistic
analytical models, static and dynamic testing of bridge components and foundation
elements should provide IDOT with a clearer picture of how to refine its ERS
strategy. The aspects of the strategy that are probably somewhat more conservative
than required should be able to be refined because the uncertainty surrounding thedifferences between actual behavior under seismic loading and that assumed in thecurrent simplified design models can be better quantified.
Stage 1: Refinement and Calibration of Seismic Structural Redundancy Level 1
A series of laboratory tests are in the planning stages, and computational
simulations are ongoing. The tests and simulations are planned to document theforce-deformation relationship, cyclic energy dissipation, and deformation capacity
during loading of bearing assemblies that are commonly used in the areas of Illinois
which are prone to seismic activity. These bearings are intended to constitute a
seismic fuse, so they are desired to be the first primary elements to fail (i.e. aresacrificial elements) within the structural system. Horizontal shearing combined with339

the effects from gravity load on side retainers and other bearing elements are underinvestigation. The experimental and computational results are expected to produce
better quantified “fuse capacities”, as well as provide information for assessing the
response of the superstructures and substructures as the fuses breach their capacities.
The computational work in this stage includes a two-tiered analysis approachfor simulating the local structural response of bridge bearing fuse assemblies. The
first set of analyses is using component modeling concepts to characterize the likely
range of response of the bridge bearings. The formulation utilizes a combination ofcoarse-mesh continuum elements and beam or spring elements that allows forparametric bracketing of the likely variables which govern the response of bridge
bearing assemblies and neighboring components of a bridge’s superstructure. These
parametric studies are helping to highlight the details of what needs to be assessed in
the experimental study. The second set are more detailed continuum type analyses of
the test specimens that allow for both prediction and corroboration of the experiments,
and to enable exploration of a wide range of parametric variables than can be assessedin the tests. Both types of analyses include nonlinear constitutive response of the
bearing and, as needed, the bridge superstructure, geometric nonlinearity based on the
anticipated deformations, and gap/contact/friction response (where appropriate).
The experiments in this research stage are planned to focus on the three bridge
bearing types most commonly used in current IDOT design and construction practice:
1. Standard Illinois “Low-profile” fixed bearings
2. Standard Illinois “Type I” steel reinforced elastomeric expansion bearings
3. Standard Illinois “Type II” steel reinforced elastomeric expansion bearings
with a slider surface
All bearing types are planned to be experimentally evaluated at full-scale forlongitudinal, transverse and bi-directional (skew) horizontal loading conditions. Figs.
6, 7 and 8 illustrate the preliminary rig configuration for the testing of bearing fusecapacities at the University of Illinois’ Newmark Structural Engineering Laboratory.
Fig. 6 presents a plan view, and Figs. 7 and 8 present elevation views.
Stage 2: Computational Simulation of Response of Bridge Systems
Based on calibrations of the fuse responses determined from Stage 1, a seriesof parametric computational analyses are planned for a suite of typical bridge systems
in order to investigate the response of the superstructure and substructure toappropriate seismic excitations. Specific issues to be addressed in these analyses
include: a) Documentation of the progression of damage (sequence of fusing) in thebridge, to ensure that proper fuses typically fail first (rather than other portions of thesuperstructure or substructure); b) Investigation of the required seat widths to ensure
adequate bridge performance; c) Documentation of anticipated peak forces that willbe transmitted to the substructure, and; d) Evaluation of changes in stiffness orstrength characteristics (such as period) after a seismic event. Fig. 9 illustrates atypical preliminary bridge system model. Parameters that will be varied in the global

340

bridge system modeling include: bearing type; superstructure type (steel girder andprestressed concrete girder); single vs. multi-span (with the strong focus being
especially on continuous multi-span systems); and selected, common substructure
types (including pier and wall assemblies, integral abutments, appropriate soil andembankment conditions, and related parameters). Of particular interest in thesesystem studies is addressing concern about vertical accelerations unseating the girders
on the pintles of low profile fixed bearings in construction typical to Illinois, as wellas assessing appropriate seat-widths for damage due to dynamic loading.

FIG. 6. PRELIMINARY PLAN VIEW OF BEARING TEST SETUP

FIG. 7. PRELIMINARY ELEVATION VIEW #1 OF BEARING TEST SETUP

341

FIG. 8. PRELIMINARY ELEVATION VIEW #2 OF BEARING TEST SETUP

FIG. 9. PRELIMINARY TYPICAL SYSTEM ANALYSIS MODEL
The modeling strategies will typically include detailed macro-level component
models of the bridge, where the girders, deck, bearings, and substructure components
are modeled explicitly. Model components outside the scope of testing in this research
will be calibrated in part based on existing seismic experimental studies of appropriatecomponents available in the literature. Both equivalent static and dynamic analyses
will be included in these studies. The equivalent static (pushover) will be used toassess monotonic strengths (system capacity), whereas the dynamic analyses will be
used to corroborate the monotonic strengths as well as to determine estimated seismic
demands. Seismic records appropriate for Mid-America have been reported in theliterature and will be utilized for this study.

342

Stage 3: Refinement of Strength Reduction Factors (and R-factors

Plans for the final stage of the project include adequately processing thesystem analyses from Stage 2 in order to assess appropriate and calibrated seismic
strength reduction factors ( and design values for R-factors, and to begin
development of an appropriate simplified method for pushover analysis for use as partof the typical design procedure in highly seismic regions of Illinois. Simplified
pushover analysis in the LRFD Seismic Guide Specifications is not well developed formany typical bridge systems used in Illinois.
Acknowledgements
The funding for the ongoing research project described in this paper, ICT R2770: Calibration and Refinement of Illinois’ Earthquake Resisting System Bridge
Design Methodology, provided by the Illinois Center for Transportation (ICT), is
gratefully acknowledged.
The authors also acknowledge the contributions of Mr. Thomas J.
Domagalski, Mr. Michael A. Mossman, Mr. Greg D. Farmer, and Ms. Rebecca M.
Leach. Mr. Domagalski, Unit Chief of Design in the Bureau of Bridges andStructures at IDOT, contributed valuable advice and guidance during development of
the IDOT ERS strategy. Mr. Mossman, Mr. Farmer, and Ms. Leach produced thedetailed drawings in the initial part of this paper.
References
American Association of State Highway and Transportation Officials (AASHTO).
(2000). Guide specifications for seismic isolation design with interims,
Washington, D.C.
AASHTO. (2002). Standard specifications for highway bridges, Washington, D.C.
AASHTO. (2009a). Guide specifications for LRFD seismic bridge design,
Washington, D.C.
AASHTO. (2009b). LRFD bridge design specifications and interim specifications,
Washington, D.C.
Building Seismic Safety Council (BSSC). (1995). “1994 edition NEHRP
recommended provisions for seismic regulations for new buildings and otherstructures.” Rep. Nos. FEMA-222a and FEMA-223a, prepared by BSSC for
Federal Emergency Management Agency (FEMA), Washington, D.C.
Federal Emergency Management Agency (FEMA). (1988). “NEHRP recommended
provisions for the development of seismic regulations for new buildings.” Rep.
Nos. FEMA-95 Part 1: Provisions and FEMA-96 Part 2: Commentary,
FEMA, Washington, D.C.

343

Federal Highway Administration (FHWA). (2006). “Seismic retrofitting manual for
highway structures, Part 1 – Bridges.” Pub. No. FHWA-HRT-06-032, FHWA,
Washington, D.C.
Hodel, C. E., Trello, M. J., and Tobias, D. H. (2004). “Seismic retrofitting in Illinois:
bridge population implementation.” Proc., 4th Nat. Seismic Conf. andWorkshop on Bridges and Highways, Federal Highway Administration,
Washington, D.C.
Illinois Department of Transportation (IDOT). (2008a). Bridge Manual, Springfield,
IL.
IDOT. (2008b). Seismic Design Guide, Springfield, IL.
International Code Council (ICC). (2000). International Building Code, Birmingham,
AL.
Leyendecker, E., Frankel, A., and Rukstales, K. (2007). “AASHTO guidespecifications for LRFD seismic bridge design.” AASHTO, Washington, D.C.
Multidisciplinary Center for Earthquake Engineering Research (MCEER)/Applied
Technology Council(ATC). (2003). “Recommended LRFD guidelines for theseismic design of highway bridges.” Special Pub. No. MCEER-03-SP03,
MCEER, Buffalo, NY.
National Cooperative Highway Research Program (NCHRP). (2006). “Recommended
LRFD guidelines for the seismic design of highway bridges.” Draft Rep.
NCHRP 20-07 Task 193, TRC Imbsen and Associates, Sacramento, CA.
Tobias, D. H., Anderson, R. E., Hodel, C. E., Kramer, W. M., and Wahab, R. M.
(2006a). “Implications of future seismic bridge design provisions for Illinois.”
Proc., 5th Nat. Seismic Conf. and Workshop on Bridges and Highways, Federal
Highway Administration, Washington, D.C.
Tobias, D. H., Anderson, R. E., Kramer, W. M., Hodel, C. E., and Wahab, R. M.
(2006b). “Standardized seismic bridge design in Illinois.” Proc., ASCE
Structures Congress 2006, American Society of Civil Engineers, Reston, VA,
1-8.
Tobias, D. H., Anderson, R. E., Hodel, C. E., Kramer, W. M., Wahab, R. M., andChaput, R. J. (2008a). “Implementation of 2008 AASHTO Seismic Bridge
Design Provisions for Illinois.” Proc., 6th Nat. Seismic Conf. and Workshop on
Bridges and Highways, Federal Highway Administration, Washington, D.C.
Tobias, D. H., Anderson, R. E., Hodel, C. E., Kramer, W. M., Wahab, R. M., andChaput, R. J. (2008b). “Overview of Earthquake Resisting System Design andRetrofit Strategy for Bridges in Illinois.” Pract. Per. on Structural Design andConstr., American Society of Civil Engineers, Reston, VA, Vol. 13, No. 3,
147-158.

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25th US-Japan
Bridge Engineering Workshop

Resolution

Page Left Blank

RESOLUTIONS
OF THE TWENTY-FIFTH U.S.-JAPAN BRIDGE ENGINEERINGWORKSHOP (Draft)
TASK COMMITTEE G, TRANSPORTATION SYSTEM
U.S.-JAPAN PANAL ON WIND AND SEISMIC EFFECTS (UJNR)
Tsukuba, Japan
19-21 October 2009

The following resolutions are hereby adopted:
1.

The Twenty-fifth U.S.-Japan Bridge Engineering Workshop provided
valuable exchange of technical information that was beneficial to bothcountries. Thirteen U.S. side participants and thirty-one Japan-side
participants discussed the subjects of Seismic Engineering andManagement of highway bridges. In view of the importance ofcooperative programs on the subjects of bridge engineering, thecontinuation of the Workshop is considered essential.

2.

Because of the importance and benefits of the Workshop, it isrecommended to hold the Twenty-sixth U.S.-Japan Bridge Engineering
Workshop in the autumn of 2010 in the United States. Dates, location,
themes, program and itinerary will be proposed by the U.S.-side Task
Committee G with the concurrence of the Japan-side Task Committee G.

347

Page Left Blank

25th US-Japan
Bridge Engineering Workshop

Appendix 1:
Japan Participants
Taro Awamoto
Taku Hanai
Jun-ichi Hoshikuma
Koichi Ikuta
Kazuhiko Kawashima
Takayoshi Kodama
Yuji Mishima
Chiaki Nagao
Takao Okada
Shinya Omote

Junichi Sakai
Masahiro Sirato
Yoshihiko Taira
Yoshiki Tanaka
Hiroshi Watanabe
Eiki Yamaguchi
Koichi Yokoyama
Atsushi Yoshioka
Tsutomu Yoshioka
Shinsuke Yumoto
Guangfeng Zhang

Page Left Blank

Taro Awamoto
Senior Staff Member
Maintenance Section
Road and Street Administration Division
Bureau of Construction
Tokyo Metropolitan Government
2-8-1 Nishi-shinjuku Shinjuku-ku Tokyo, 163-8001, Japan
Phone: +81-3-5320-5366
Fax:+81-3-5388-1529
E-mail:Tarou_Awamoto@member.metro.tokyo.jp

Taku HanaiI
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: hanai@pwri.go.jp

Jun-ichi Hoshikuma, Dr.Eng.
Chief Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6793
Fax:+81-29-879-6739
E-mail: hosikuma@pwri.go.jp

Koichi Ikuta
Researcher
Bridge and Structures Division, Road Department,
National Institute for Land and Infrastructure Management
Ministry of Land, Infrastructure, Transport and Tourism
1 Asahi, Tsukuba, Ibaraki, 305-0804 Japan
Phone: +81-29-864-4919
Fax:+81-29-864-4917
E-mail: ikuta-k8810@nilim.go.jp351

Kazuhiko Kawashima, Dr.Eng.
Professor
Department of Civil Engineering
Tokyo Institute of Technology
M1-10, O-Okayama,Meguro, Tokyo, 152-8552, Japan
Phone: +81-3-5734-2922
Fax:+81-3-5734-3810
E-mail: kawashima.k.ae@m.titech.ac.jp

Takayoshi Kodama
Manager of Technology Department
Kajima road company, Tokyo
7-27, 1cho-me Kouraku, Bunkyo-city, Tokyo, Japan
Phone: +81-3-5802-8014
Fax:+81-3-5802-8045
E-mail: kodama@kajimaroad.co.jp

Yuji Mishima
Manager, Infrastructure Technology Development Dept.,
Hitachi Zosen Corporation
7-89, Nanko-kita 1-chome, Suminoe-ku, Osaka, Japan
Phone: +81-6-6569-0051
Fax:+81-6-6569-0055
E-mail: mishima_y@hitachizosen.co.jp

Chiaki Nagao
Engineer
Central Nippon Expressway Company Limited
Planning and Design Team Nagoya Branch Central
Nippon Expressway Company Limited;
Mitsui-Sumitomo Bank Nagoya Building, 2-18-19, Nishiki, Naka-ku, Nagoya, 460-0003
Phone: +81-52-222-1594
Fax:+81-52-232-3718
E-mail: c.nagao.aa@c-nexco.co.jp

Takao Okada
Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: okada@pwri.go.jp352

Shin-ya Omote
Researcher
Structures Research Team
Civil Engineering Research Institute for Cold Region
Public Works Research Institute
Hiragishi 1-3-1-34, Toyohira-ku, Sapporo, 062-8602, Japan
Phone: +81-11-841-1698
Fax:+81-11-841-3502
E-mail: omote@ceri.go.jp

Junichi Sakai, Dr.Eng.
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: sakai55@pwri.go.jp

Masahiro Shirato, Dr.Eng.
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: shirato@pwri.go.jp

Yoshihiko Taira
Senior Structural Engineer
Sumitomo Mitsui Construction Co., Ltd.
1-38-1 Chuo Nakano-ku, Tokyo, 164-0011, Japan
Phone: +81-3-5337-2134
Fax:+81-6-3367-4763
E-mail: ytaira@smcon.co.jp

Yoshiki Tanaka
Senior Research Engineer
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: tanakay@pwri.go.jp353

Hiroshi Watanabe
Leader of Construction Material Team
Material and Geotechnical Engineering Research Group
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6760
Fax:+81-29-879-6736
E-mail: hwatana@pwri.go.jp

Eiki Yamaguchi, Ph.D.
Professor of Civil Engineering
Dept. of Civil Engineering
Kyushu Institute of Technology
Kitakyushu, 804-8550, Japan
Phone: +81-93-884-3110
Fax:+81-93-884-3100
E-mail: yamaguch@civil.kyutech.ac.jp

Atsushi Yoshioka
Director
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6726
Fax:+81-29-879-6739
E-mail: yoshioka@pwri.go.jp

Tsutomu Yoshioka
Chief of Maintenance Engineering
Nippon Engineering Consultants Co., Ltd
3-23-1 Komagome Toshimaku Tokyo, Japan, 170-0003
Phone: +81-3-5394-7604
Fax:+81-3-5394-7606
E-mail: yoshioka@ne-con.co.jp

Shinsuke Yumoto
Staff of Bridges Department
Public Works Division, Public Works Bureau
Osaka City Government
WTC-building 12F,1-14-16 Nankokita Suminoe-Ku Osaka City, 559-0034,Japan
Phone: +81-6-6615-6818
Fax:+81-6-6615-6583
E-mail: s-yumoto@city.osaka.lg.jp354

Guangfeng Zhang
Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: tyou44@pwri.go.jp355

blank25th US-Japan
Bridge Engineering Workshop

Appendix 2:
US Participants
Abdeldjelil "DJ" Belarbi
X. Hannah Cheng
Catherine French
Jugesh Kapur
Sashi Kunnath
Manos Maragakis
Barney T. MartinJr.

Barton Newton
M. “Saiid” Saiidi
David H. Sanders
Daniel H. Tobias
Bojidar Yanev
W. Phillip Yen

Page Left Blank

Abdeldjelil "DJ" Belarbi
Distinguished Professor
Department of Civil, Architectural and Environmental Engineering
Missouri University of Science and Technology (Missouri S&T)
323 Butler-Carlton Hall 1401 North Pine Street Rolla, MO 65409-0030
Phone: 573-341-4478
Fax: 573-202-2117
E-mail: belarbi@mst.edu

X. Hannah Cheng
Civil Engineer, Bureau of Structural Engineering,
New Jersey Department of Transportation
1035 Parkway Ave., Trenton, NJ 08625
Phone: 609- 530-2464
Fax: 609-530-5777
E-mail: Xiaohua.Cheng@dot.state.nj.us

Catherine FrenchInstitute of Technology Distinguished Professor Department of Civil
Engineering University of Minnesota
122 CivE Bldg.
500 Pillsbury Dr. S.E.
Minneapolis, MN 55455-0220
Phone: 612-625-3877
Fax: 612-626-7750
E-mail: cfrench@umn.edu

Jugesh Kapur
Washington State DOT
State Bridge & Structures Engineer
WSDOT Bridge and Structures Office
P O Box 47340 Olympia, WA 98504-7340
Phone: 360-705-7207
Fax: 360-705-6814
E-mail: KapurJu@wsdot.wa.gov

Sashi Kunnath
Professor & Chair
Department of Civil & Environmental Engineering
University of California
Davis CA 95616
Phone: 530-754-6428
Fax: 530-752-7872
E-mail: skkunnath@ucdavis.edu

359

Barney T. Martin, Jr.
President
Modjeski and Masters, Inc.
301 Manchester Road, Suite 102
Poughkeepsie, New York 12603
Phone: 845-471-2630
Fax: 845-471-4259
E-mail: btmartin@modjeski.com

Barton Newton
State Bridge Maintenance Engineer
California Department of Transportation (Caltrans)
1801 30th Street, MS 9-1/9i
Sacramento, California 95816
Phone: 916-227-8841
Fax: 916-227-8357
E-mail: barton.newton@dot.ca.gov

M. “Saiid” Saiidi
Professor
Civil and Environmental Engineering Department (258)
University of Nevada
Reno, NV 89509
Phone: 775-784-4839
Fax: 775-784-1390
E-mail: Saiidi@unr.edu

David H. Sanders
Professor
Department of Civil and Environmental Engineering
University of Nevada, Reno
1664 North Virginia Street-MS 258
Reno, NV 89557-0258
Phone: 775-784-4288
Fax: 775-784-1390
E-mail: sanders@unr.edu

Daniel H. Tobias
Bureau of Materials and Physical Research
Illinois Department of Transportation
126 East Ash Street
Springfield, IL 62704-4766
Phone: 217-782-2912
Fax: 217-782-2572
E-Mail: Daniel.Tobias@illinois.gov

360Bojidar Yanev
Executive Director, Bridge Inspection & Management,
Department of Transportation, New York City
55 Water St., New York, NY10038
Phone: 212 839 4181
Fax:
E-mail: byanev@dot.nyc.gov

W. Phillip Yen
Seismic Research Program Manager
Office of Infrastructure, R&D
Federal Highway Administration
6300 Georgetown Pike
McLean, VA 22101
Phone: 202-493-3056
Fax: 202-493-3442
E-mail: Wen-huei.Yen@fhwa.dot.gov

361