ISSN 0386-5878
Technical Reportof PWRI No.4218

PROCEEDINGS OF
THE 27th U.S. – JAPANBRIDGE ENGINEERING WORKSHOP
Tsukuba, Japan2011
November 7, 8 and 9, 20

CENTER FOR ADVANCED ENGINEERING STRUCTUR
STRUCTURAL
AL ASSESSMENT AND RESEARCHRE SEARCH

PUBLIC WORKS RESEARCHSEARCH INSTITUTE
Incorporated Administrative Agency

1-6,
6, Minamihara, Tsukuba
Tsukuba-shi, Ibaraki-ken,
ken, 305
305-8516
8516 Japan

Opinions, findings, conclusions and recommendations expressed in the papers contained inthis publication are those of the author(s).

Copyright © (2012) by P.W.R.I.
All rights reserved.

No part of this book may be reproduced by anymeans, nor transmitted, nor translated into a machine language without
the written permission of the Chief Executive of P.W.R.I.

ࡇ ࡢሗ ࿌᭩ ࡣࠊ ⊂ ❧⾜ ᨻἲ ேᅵ ᮌ◊ ✲ᡤ ⌮ ஦㛗 ࡢᢎ ㄆࢆ ᚓ࡚ ห⾜ ࡋࡓࡶࡢ࡛࠶ࡿࠋࡋࡓࡀࡗ࡚ࠊᮏሗ࿌᭩ࡢ඲㒊ཪࡣ୍㒊ࡢ㌿㍕ࠊ」〇ࡣࠊ⊂❧ ⾜ᨻ ἲே ᅵᮌ ◊ ✲ᡤ ⌮஦ 㛗ࡢ ᩥ᭩ ࡟ࡼ ࡿ ᢎㄆ ࢆᚓ ࡎࡋ ࡚ࡇ ࢀࢆ ⾜ࡗ࡚ࡣ࡞ࡽ࡞࠸ࠋ

Technical Reportof PWRI No.4218

PROCEEDINGS OF
THE 27th U.S. – JAPANBRIDGE ENGINEERING WORKSHOP

By
Editor: Hideaki Nishida

Synopsis:
The proceeding documents the results of the 27th U.S. – Japan Bridge Engineering Workshop
which was held at Tsukuba, Japan, on November 7, 8, and 9, 2011, as a part of the activities of
the Panel on Wind and Seismic Effects, UJNR. The Workshop was organized by Task Committee
G “Transportation System” (U.S. side chair: Dr. W. Philip Yen, FHWA, Japan side chair: Mr.
Tetsurou Kuwabara, CAESAR, PWRI) of the Panel.

Key Words: Bridge, Design, Construction, Maintenance, Seismic, UJNRPage Left Blank

The steering committee for the 27th US-Japan Bridge Engineering Workshop
consisted of Tetsurou Kuwabara, Phillip Yen, David Sanders, and Hideaki Nishida.
The workshop was held at the NILIM and the CAESAR, PWRI, in Tsukuba, Japan.
The 2-1/2 day workshop focused on: 1) Great East Japan Earthquake, 2) Tsunami 3)
Load and Strength Evaluation, 4) Maintenance, 5) Inspection, 6) Design andAnalysis, and 7) Acceleration Bridge Construction. Thirteen participants from the
US and thirty-seven participants from Japan attended the workshop who werearranged by both T/C chairs in terms of the focused themes. The papers contained
within this proceeding are the papers that were presented at the workshop (17papers from the Japan side and 13 papers from the US side).
In addition to the workshop, there was a bridge study held after the workshop,
November 10-12, 2011, visiting bridge sites:-

Long-term Bridge Management Plan for Old Bridges, Kiyosu Bridge and Eitai
Bridge (Tokyo Metropolitan Government)-

Improvementof

Seismic

Performance,

Katsushima

area(Metropolitan

Expressway)-

Damaged Bridges due to Great East Japan Earthquake(a)Arakawa Wangan Bridge (Metropritan Expressway)(b)Koizumi Oh-hashi Bridge,Kesen Oh-hashi Bridge ,Osaragi Bridge andKameda Oh-hashi Bridge (Tohoku Regional Development Bureau ,MLIT)(c)Kamata Oh-hashi Bridge (Fukushima city)in the Kanto and Tohoku regions.
ACKNOWLEDGEMENTS
There are many people that made the 27th US-Japan Bridge Engineering Workshop
a success. I really appreciate cooperation of Tokyo Metropolitan Government,
Metropolitan Expressway, Tohoku Regional Development Bureau and Ibaraki
Prefecture.
Editor: Hideaki Nishida, Senior Researcher, Bridges and Structures Research Group,Center for Advanced Engineering Structural Assessment and Research (CAESAR),
Public Works Research Institute (PWRI)TABLE OF CONTENTS
27 US-Japan Bridge Engineering Workshop
Tsukuba, Ibaraki November 7 to 9, 2011th

PREFACETABLE OF CONTENTS
WORKSHOP PROGRAM············································································ 1
SESSION 1: Great East Japan Earthquake······················································· 9
1-1

Damage of Highway Bridges due to the Great East Japan Earthquake
By Junichi Hoshikuma ····································································· 11

1-2

Ground Motion and Tsunami Induced Damage of Bridges during 2011 Great East
Japan Earthquake
By Kazuhiko Kawashima, Kenji Kosa, Yoshikazu Takahashi, Mitsuyoshi Akiyama,and Hiroshi Matsuzaki ····································································· 23

1-3

Bridge Performance during the Great East Japan Earthquake of March 11, 2011
By M. Lee Marsh, Ian G. Buckle, W. Phillip Yen, Shideh Dasti, David Frost,and Eric Monzon············································································ 39

SESSION 2: Tsunami and Seismic1 ······························································ 53
2-1

Damage Analysis of Bridges Affected by Tsunami due to Great East Japan
Earthquake
By Kenji Kosa ·············································································· 55

2-2

Development of Tsunami Design Criteria for Oregon Coastal Bridges
By Bruce Johnson and Solomon Yim····················································· 67

2-3

FHWA Seismic Hazard Mitigation Research Program
By W. Phillip Yen ········································································· 83iSESSION 3: Seismic2 ··············································································· 91
3-1

Failure Investigation and Retrofit Strategies of a 22-Span Steel Girder
Bridge during the 2010 Chile Earthquake
By Genda Chen, Zuocai Wang, Phillip W. Yen, and Ian Buckle ······················ 93

3-2

Earthquake Response of Highway Bridges Subjected to Long Duration Seismic
Motions
By Shojiro Kataoka········································································ 107

3-3

Experimental Investigation of Influence of Live Load on Seismic Response of a
Horizontally Curved Bridge
By Ian G. Buckle, Hartanto Wibowo, Danielle M. Smith,and David H. Sanders ····································································· 119

3-4

Discussion Method by Using Dynamic Analysis for Quakeproofing Long-Span
Bridges
By Shunji Nagahashi, Takumi Nishikawa, Mutsuji Kawakami,
and Tatsumi Ujie ········································································· 133

SESSION 4: Management and Maintenance1 ················································· 143
4-1

Analysis Examples of Periodic Inspection Results of National Highway Bridges
in Japan-Application of Degree of Damage for Element Level-

By Fumihiko Nomura, Masanori Okubo, and Takashi Tamakoshi ··················· 145
4-2

The Importance of Partnerships in the Implementation of Bridge
Preservation Practices
By Peter J. Weykamp······································································ 157

4-3

Maintenance of ASR-Affected Structures in Hokuriku Expressway, JapanBy Yoshifumi Adachi, Mamoru Moriyama, and Masahiro Nomura ················· 167

SESSION 5: Load and Strength Evaluation ··················································· 175
5-1

Load-Carrying Capacity of Reinforced Concrete Beams with Adhesively Bonded
Steel Plates
By Yoshiki Tanaka, Jun Murakoshi, and Eiji Yoshida ································· 177
ii

5-2

Summary of NCHRP Research on Recalibration of the LRFR Load Factors in the
AASHTO Manual for Bridge Evaluation
By Waseem Dekelbab ····································································· 191

5-3

Compressive Loading Test of Corroded Gusset Plate Connection in Steel Truss
BridgeBy Jun Murakoshi, Naoki Toyama, Mamoru Sawada, Kentaro Arimura,
Lu Guo, Kuniei Nogami, Teruhiko Yoda, and Hideyuki Kasano ····················· 201

SESSION 6: Maintenance2 ······································································· 219
6-1

Introduction of Non-Destructive Highway Inspection Methods Using HighDefinition Video and Infrared Technology
By Koji Mitani and Masato Matsumoto················································· 221

6-2

Charpy Impact Tests with Test Specimens Made with
Stop-Hole-Size Cores
By Kiyoshi Ono, Kengo Anami, and Mitsuharu Oikawa······························ 233

6-3

The Next Generation of Experimental Research in Structural Fires
By John L. Gross and Stephen A. Cauffman ··········································· 243

6-4

New Attempt to Maintenance of Steel Bridge Coating in Expressway
By Shuhei Sakai and Dai Wakabayashi ················································· 255

6-5

FHWA’s New Infrastructure Research and Technology Strategic PlanBy Ian M. Friedland ······································································· 267

SESSION 7: Design and Analysis ································································ 279
7-1

Unbonded Prestressed Columns for Earthquake Resistance
By David H. Sanders, Alex S. Larkin, and M. Saiid Saiidi ··························· 281

7-2

Study on Corrosion Analysis with Fiber Model of Long-Span Truss Bridge
By Kunihiro Hayashi, Hideo Ohishi, Yukio Adachi, Misa Fujibayashi,
and Koichi Sugioka········································································ 291
iii

7-3

Seismic Performance Assessment of Concrete Bridges Designed by
Displacement-Based Methods
By Kevin Mackie, Vinicio Suarez, and Oh-Sung Kwon ······························ 301

7-4

Local Buckling Analysis of Steel Truss Bridge under Seismic Loading
By Eiki Yamaguchi and Keita Yamada ·················································· 315

SESSION 8: Construction and Maintenance3 ················································· 323
8-1

Use of Precast I-Girders for Accelerated Bridge Construction in High Seismic
RegionsBy Sri Sritharan, Justin Vander Werff, and Rick Snyder ······························ 325

8-2

How to Build a Bridge – Fast
By John Stanton············································································ 337

8-3

Research and Repair of Reinforcement by Overlaying Floor Slab Method in
Expressway
By Hidekazu Hayashi ····································································· 349

8-4

Repairing Fatigue Cracks in a Steel Deck Plate Girder Bridgeof Unusual Structure
By Haruhiko Nakata, Yoshiyuki Takamura, and Ken Tokumasu ····················· 359

8-5Development of Eddy Current Test to Detect Fatigue Damages
on Orthotropic Steel Deck
By Hiroki Sugiyama, Akiko Tabata, Hiroyuki Nakajima, Tetsuji Yamagami,
Sigeaki Tsukamoto, Akira Shiraishi, and Takaaki Yamada ··························· 373

APPENDIX 1 : JAPAN PARTICIPANTS ······················································· 383
Resume of Japan Participants
APPENDIX 2 : US PARTICIPANTS ···························································· 393
Resume of U.S. Participantsiv

27th US-Japan
Bridge Engineering Workshop

Workshop ProgramPage Left Blank

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䋭8䋭27th US-Japan
Bridge Engineering Workshop

Session

1

Great East Japan Earthquake
Damage of Highway Bridges Due to the Great East Japan Earthquake
By Junichi Hoshikuma
Ground Motion and Tsunami Induced Damage of Bridges during 2011 Great East Japan
Earthquake
By Kazuhiko Kawashima, Kenji Kosa, Yoshikazu Takahashi, Mitsuyoshi Akiyama,and Hiroshi Matsuzaki
Bridge Performance during the Great East Japan Earthquake of March 11, 2011
By M. Lee Marsh, Ian G. Buckle, W. Phillip Yen, Shideh Dasti, David Frost,and Eric Monzon

Page Left Blank

䋭10䋭

DAMAGE OF HIGHWY BRIDGES DUE TO
THE GREAT EAST JAPAN EARTHQUAKE
Jun-ichi Hoshikuma 1
Abstract
The 2011 Great East Japan Earthquake struck the Tohoku and Kanto area on
March 11, 2011. Immediately after the earthquake, research engineers in NILIM/PWRI
were dispatched to Tohoku area to investigate the damage to bridges and suggest thetechnical advice to bridge administrators. On behalf of NILIM/PWRI jointreconnaissance team for bridge damage, author summarizes damage characteristic ofbridges due to the earthquake in this paper. Technical issues learned from the damageof bridges were also described based on the investigation of bridge damage.
Introduction
The 2011 Great East Japan Earthquake occurred at 2:46 pm on March 11, 2011.
The catastrophic damage resulting from strong ground motion and huge tsunami wascaused in Tohoku and Kanto regions. More than 20,000 people were killed or missing
and various infrastructures were damaged, especially in the coastal area of Iwate,
Miyagi, Fukushima and Ibaraki Prefectures.
Many highway bridges were also damaged in these areas due to both large
ground motion and tsunami inundation. Soon after the earthquake occurred, NILIM and
CAESAR in PWRI jointly investigated bridge damage and provided the technicalsupports and suggestions to bridge administrators, including Regional Bureaus of
MLIT and some Local Governments. Furthermore, Task Committee G of UJNR Panel
conducted the U.S.-Japan joint reconnaissance to bridge damage in early June, 2011.
This paper presents damage characteristic of bridges. Following topics werefocused; bridge damage due to tsunami inundation or strong ground motion effect,
verification of seismic performance of bridges retrofitted after the 1995 Hyogo-ken
Nambu earthquake, validations of effectiveness of the current seismic design
specification.
The 2011 Great East Japan Earthquake and Ground Motion
The main shock of this earthquake (Mw=9.0, focal depth=24km) occurred at
2:46 pm (JST) on March 11, 2011. Maximum seismic intensity was observed at
Tsukidate, Kurihara city in Miyagi prefecture (Seismic intensity of JMA was 7) andlarge seismic intensities were observed in Tohoku and Kanto areas. Fig. 1 shows
acceleration ground motion waveforms and spectral response accelerations at
representative strong ground motion observation sites.

1

Chief Researcher, Center for Advanced Engineering Structural Assessment andResearch, PWRI

䋭11䋭

Fig. 1 Acceleration Waveforms and Spectral Response Acceleration at Main Shock(NS comp.)

䋭12䋭

It should be noted that 1) strong ground motion records with long duration wereobserved and 2) there were multiple pulses in some ground motion records observed
near epicenter. This is because large fault areas collapsed continuously. It was observed
at very large maximum response acceleration at the range of short predominant period
such as Tsukidate record. The maximum response accelerations at the range of natural
periods from 1.0 to 2.0 seconds, which relatively correlate with damage of ordinary
road bridges, were equal or slightly less than those of the 1995 Hyogo-ken Nambu
earthquake. Ground motions and maximum response accelerations at the coastal areaof Tohoku region were not so large. However, strong ground motions and large
response accelerations were observed at the sites where located slightly far from
epicenter such as Fukushima, Tochigi and Ibaraki prefectures.
Huge tsunami induced by main shock struck at Tohoku and Kanto coastal areas
and exceeding 10m in height of wave were observed.
Moreover, aftershocks with the JMA magnitude of 7.0 or over were occurred
three times within a day and total of 89 aftershocks with the magnitude of 6.0 or overwere occurred until August 3.
Overview of Damage in Bridges
Damage of the highway bridges due to this earthquake can be categorized as
effect of strong ground motion, effect of tsunami inundation, and effect of soilliquefaction. It should be noted in this earthquake that the intensive damage in highway
bridges was mainly caused by tsunami inundation. Superstructures in twelve bridges
including service road for pedestrian on national highway route 45 (main route along
the Pacific coast of Tohoku Area) were washed away, which resulted in the traffic close
after the earthquake. About 91 highway bridges in total were washed away due totsunami inundation in Iwate, Miyagi, Fukushima, Ibaraki and Chiba prefectures. On theother hand, as long as we have investigated, 105 bridges survived even though thesuperstructures of these bridges were inundated with the tsunami. The backfill of
abutment in some bridges were also washed out even though super- and substructures
survived.
The ground motion effect to damage of bridges was less significant than thetsunami effect. One bridge (Rokko Ohashi Bridge, an old steel girder bridge supported
by steel pile-bent columns located in Ibaraki prefecture) was collapsed due to theground motion of the earthquake. Although the collapsed bridge was observed at the
only Rokko Ohashi in the highway bridges, it was found in the bridge designed inaccordance with pre-1980 design specifications that damage to RC columns at sectionof cut-off of longitudinal rebars, damage to RC pier-wall with small amount ofreinforcement, damage to steel bearings and attachment of bearings, damage to bracing
and steel members, and subsidence of backfill soil of abutment. These damage modes
have already observed in the past earthquakes. However, rupture of elastometric rubber
bearings were observed at the Sendai-Tohbu viaduct designed based on Post-1995
design specifications.
After the Kobe Earthquake, the seismic retrofit project has been performed forexisting bridges columns designed in accordance with pre-1980 specifications withhigh priority, to prevent the collapse of the bridge structure and unseating of the deck.

䋭13䋭

Almost of retrofitted bridge columns were not damaged due to the ground motion of theearthquake, which would exhibit the effectiveness of the seismic retrofit.
Soil liquefaction was widely observed in particularly Tokyo Bay area. Although
the effect of the soil liquefaction on the bridge damage was minor, subsidence of
backfill soil of abutment due to the soil liquefaction effect was developed in somebridges. Deck-end gap was shortened resulting from movement of substructure, whichcaused steel bearings damage and cracks in parapet wall.
Bridge Damage Due to Ground Motion
Damage of Unretrofitted Bridges Designed in Accordance with Pre-1980 Design
Specifications
Intensive damage due to the ground motion was developed in many
unretrofitted bridges designed in accordance with pre-1980 design specifications.
Almost of damage modes of those bridges have ever been observed in the pastearthquakes. Photos 1 to 3 show the damage to reinforced concrete piers, steel bearing
supports and the attachment of the bearing support to pier top or the superstructure,
respectively. Although some bridge columns collapsed during the 1995 Kobe
earthquake, the only one bridge collapsed during the 2011 Great East Japan earthquake.
As shown in Photo 4, an old steel girder bridge supported by steel pile-bent columns(Rokko Ohashi Bridge) in the local roadway was collapsed due to the ground motion of
the earthquake. Rokko Ohashi is to be replaced a new bridge and the new one was being
constructed at the time of the earthquake.
Damage of Bridges Designed in Accordance with Post-1995 Design Specifications
There were few intensively-damaged bridges designed in accordance with
post-1995 design specifications, where the seismic design acceleration increased basedon the ground motion records of the 1995 Kobe earthquake and the details of thetransverse steel was specified for improving the confinement effect and shear capacityof the RC columns.

Photo 1 Damage of Reinforced Concrete Columns at Cut-off Section ofLongitudinal Reinforcement

䋭14䋭

Photo 2 Damage of Steel Bearing Support

Photo 3a Damage of Pier Top

Photo 3b Damage of Attachment of
Bearings to Superstructure

Photo 4 Collapse of Steel Pile-bent Columns

However, as shown in Photo 5, it should be noted that elastometric rubber
bearings ruptured at the Sendai-Tohbu viaduct designed based on Post-1995 design
specifications. Fig. 2 illustrates the structure of the Sendai-Tohbu viaduct and thepositions of the elastometric rubber bearing with the rupture. As seen in Fig. 2, bridge
structure in this junction is very complicate. Bridge width changes significantly in the
4-span continuous box girder with the change of the section. Span length of the section
from P52 to 56 is around 70m, while that from P56 to P58 is 39m. P52 and P53 aresingle-column hammerhead steel piers, while P54, P55 and P56 are two-column steel
frames. Rupture of the elastometric rubber bearings in the transverse direction wasobserved at the deck-end of P52 side in 4-span continuous box girder, though there arefew damaged elastometric rubber bearings at the other deck-end side of the box girder
on Pier 56. Photo 6 shows failure mode of the elastometric rubber bearing with therupture. Comparison of the failure mode with the result of the shear loading test should
be made, to estimate the capacity of the elastometric rubber bearing and the applied
seismic force combination.

䋭15䋭

Based on the lessons learned from the 1995 Kobe Earthquake, the seismic
retrofit project has been performed for existing bridges columns designed inaccordance with pre-1980 specifications with high priority, to prevent the collapse of
the bridge structure and unseating of the deck. During the 2011 Great East Japan
Earthquake, many retrofitted bridges were given a shake due to the ground motion.
Photo 7 exemplifies the effectiveness of the seismic retrofit for bridge columns.
As seen in Photo 7, there are two adjacent river-crossing bridges. Since one bridge(Nakagawa Bridge) is on the designate emergency route, bridge columns designed withthe pre-1980 specifications have already been retrofitted by reinforced concrete
jacketing. The other bridge (Kunita Ohashi Bridge) is on the local roadway and thebridge columns have not yet been retrofitted at the earthquake. Although Kunita Ohashi
Bridge suffered from the vulnerable damage and thus lost the serviceability for thebridge, Nakagawa Bridge did not suffer from the damage and kept the serviceability
soon after the earthquake. Seismic performance shown in these two bridges clearly
exhibits the effectiveness of the seismic retrofit.

Photo 7 Comparison of Seismic Performance between Adjacent Two Bridges(Nakagawa Bridge and Kunita Ohashi Bridge)
On the other hand, there are a few remarkable damage examples in the
retrofitted bridges. Photo 8 shows adjacent two reinforced concrete columns in Kameda
Ohashi Bridge. Each column supports 2-span continuous steel box girder at the middle.
The outbound column was designed with 1980 specifications and retrofitted byreinforced concrete jacketing for strengthening the cut-off section without increasing
the flexural strength of the column base. Furthermore, additional shear keys wereanchored to column top to supplement the strength of the existing bearings. Although
no damage was found to the steel bearings and shear keys, vertical cracks of nearly

䋭17䋭

10mm width were observed at the beam section as shown in Photo 8. The inboundcolumn was designed with 1994 specifications basically and some modifications weremade based on the 1995 tentative specifications (published soon after 1995 Kobe
Earthquake). In the inbound column, elastometric rubber bearings were deformed in thetransverse direction and the side stoppers were failed. Concrete of the beam edge
portion attaching the supplemental shear keys also spalled off (see Photo 8) due to thetransverse seismic force induced by the inertia of the superstructure, while verticalcrack observed in the beam of the outbound column was not developed in the inboundbeam.
Photo 8 Comparison of Damage Mode between Adjacent Two Bridge Columns(Kameda Ohashi Bridge)
Photo 9 shows the failure of pier top.
This type of damage has ever observed in
the old bridge columns since the pastearthquakes. Bearing capacity of the piertop was insufficient to transmit the seismic
force to column. It should be noted in thisbridge that the crack reaches the portion of
the anchor bars of the steel bracket for
attaching the unseating prevention devices.
Since the unseating prevention devices
should work at the unexpected situation of
failure of bearing supports.
Photo 9 Crack to Pier Top Affects Attachment

䋭18䋭Bridge Damage Due to Tsunami Inundation
Many bridges built along the coast line of the Pacific Ocean in Tohoku andKanto areas were affected by the huge tsunami. Photos 10 and 11 show the typicaldamage modes of bridge due to the tsunami inundation.
The bridge shown in Photo 10a is a 6-span steel girder bridge (twin three-spancontinuous girders), namely Koizumi Bridge. The steel girder was supported by RCpier walls with steel pipe piles. The pier walls with the fix bearings (P2 and P4) wereretrofitted by FRP sheets, furthermore viscous dampers were installed to the deck-ends.
The height of tsunami has been estimated exceeding 10 meter high in this area, whichmeans the steel girder was completely inundated with tsunami. Tsunami effect was so
significant to Koizumi Bridge that the whole girder were washed away upstream andone RC pier-wall (P3) was also washed away. The pier-wall of P3 supported twogirders with movable bearing supports was found about 50m away from original
position, while the pile foundation and the footing for P3 remained at the original
position. The pier-wall failed at the bottom section as shown in Photo 10b. The
probable reason for only P3 being washed away was due to smaller ultimate strengththan the other pier-walls. It is also noted that backfill soil of abutment was also washedaway at both sides.
Photo 11shows a 12-span PC single girder bridge, namely Utatsu Ohashi Bridge.
Piers consisted of circular RC column (P1 and P2) and rectangular RC column with PC
piles. Bridge columns were retrofitted by RC jacketing and the seat length was
extended at the top of pier. Total of 8 spans (from P2 to P10) were washed away to theinland direction. It was found that additional concrete/steel shear keys, installed at thepier top for the seismic retrofit of existing bearings, were damaged and somehammerhead beams of piers in the inland side were cracked. A lot of diagonal cracks
were also observed at both unseated and survived PC girders.
On the other hand, it is interesting to note that there are many survived bridges
even though the superstructures of these bridges must be inundated with tsunami. Photo
12 exemplifies Yanoura Bridge survived washed away of superstructure due to thetsunami, while two spans of the water pipe bridge located adjacent to Yanoura Bridge
were washed away. These survived bridges will hint the mechanism of resistance to thetsunami effect.

Impact of 2011 Great East Japan Earthquake on Seismic Design of Highway
Bridges
The seismic performance of highway bridges, designed in accordance with thepost-Kobe Japanese specifications, was very well and these bridges were functional
without any long-term traffic stops after the earthquake. However, there are several
important issues and lessons we should study and review for the latest seismic design
specifications for highway bridges. Followings are the selected issues.

䋭19䋭

Ground Motion
In the 2011 Great East Japan Earthquake, many strong ground motion records
were recorded and these records clearly showed that this earthquake generated groundmotions with multiple pulses and thus the longer duration (more than 2 minutes) thanother records observed in the past earthquakes. Similar ground motions were reported
in the 2010 Chile Earthquake with the moment magnitude Mw 8.8. Therefore, thesubduction-type earthquake with Mw of nearly 9 may induce the ground motion withlong duration.
In general, the long duration would affect the number of cyclic inelastic
response of the bridge system. Past experimental researches indicated that the loading
pattern in the quasi-static cyclic loading test, particularly the number of cyclic loading
affects the ductility capacity of flexural reinforced concrete column. In order toaccommodate such effect into the seismic design, Japanese design specifications havedetermined two ductility/shear capacity factors based on the types of the ground motion,i.e. the subduction-type and the near-fault-type. Re-studies on the effect of the long
duration will be required based on the ground motion observed in the 2011 Great East
Japan Earthquake.
The long duration would also affect the soil liquefaction. Effect of the soilliquefaction on the seismic design of bridge foundation was introduced in the 1971
specifications in Japan based on the lessons learned from the 1964 Niigata Earthquake.
Although there were no major liquefaction-induced damages in bridges during the 2011
Great East Japan Earthquake, the long duration effect on the bridge performance built
on the liquefiable sandy soil condition should be verified through both geological andstructural perspectives.
Since the ground motion effect propagated wide, bridge damage developed inwide area. Many ground motion records were also observed in wide area. It should bealso important to study the relation between the properties of the ground motion anddamage of bridges.
Tsunami Effect on Bridges
Superstructures in several bridges were washed away due to the tsunami effect.
Backfill soil for the abutment was also washed out. Similar damage modes in bridge
were also observed during the 2004 Indian Ocean Earthquake. Failure mechanism ofbridge system due to the tsunami effect need to be studied, in which the resistancecapacity of the existing bearing supports be analyzed based on both the washed-away
and survived bridges. Also, more experimental researches on the bridge behavior due tothe tsunami effect are required, to find the appropriate structural system for mitigating
the tsunami effect.
On the other hand, the design concept of bridge for unexpected extraordinary
event would be controversial, because structural resistance capacity has a limit.
Basically, it would be one of the options for the extraordinary tsunami effect to avoid
routing important highway network and locating important highway routes in thetsunami-inundation area. In terms of structural engineering, easy-to-recover bridge

䋭21䋭system employing the temporary structure with the accelerated construction technique
is also another option.
Validations of Effectiveness of Seismic Retrofit
Seismic retrofit have been performed step-by-step since 1995 Kobe earthquake.
Based on the lessons learned from the past earthquakes, bridge columns in theimportant highway network designed by pre-1980 specifications have been retrofittedwith high prioritization. Many seismic vulnerable bridges in the important route suchas National Highway Route 4, 6, 45 etc were retrofitted up to the date of the earthquake,
which resulted in quick recovery of the functional highway network after theearthquake. It should be, however, important review to investigate details of the new
type of damage in the retrofitted bridges and evaluate the seismic behavior of the bridge
during the earthquake.

Conclusion Remarks
This report preliminarily summarized damage to highway bridge due to the2011 Great East Japan Earthquake with focus of the seismic performance of theretrofitted bridges and tsunami effect. Based on the damage caused by the earthquake
and tsunami, more analytical and experimental researches should be required to clarify
the mechanism of the damage. Investigation results also indicate that subsidence of thebackfill soil in the abutment has been remarkable with the improvement of seismic
performance for bridge structures, though details are not reported in this paper. It would
be important to ensure the seismic performance of both bridge structures andembankment for highway.

Acknowledgments
Damage investigation was supported by a number of organizations. Special
appreciation is extended to Iwate Pref., Ibaraki Pref., Chiba City, Kitakami City,
Tohoku Regional Development Bureau, Kanto Regional Development Bureau andHeadquarter of MLIT.

References
1. Genda Chen, Phillip W. Yen, Ian Buckle, Tony Allen, Daniel Alzamora, Jeffrey Ger
and Juan G. Arias: 2010 Chile Earthquake Implication to the Seismic Design ofBridges, Proc. 26th US- Japan Bridge Engineering Workshop, 2010.
2. K. Kawashima, S. Unjoh, J. Hoshikuma and K. Kosa: Damage Characteristic of
Bridges due to 2010 Chile Earthquake, Proc. 26th US- Japan Bridge Engineering
Workshop, 2010.

䋭22䋭

GROUND MOTION AND TSUNAMI INDUCED DAMAGE OF BRIDGESDURING 2011 GREAT EAST JAPAN EARTHQUAKE
Kazuhiko Kawashima1, Kenji Kosa2, Yoshikazu Takahashi3, Mitsuyoshi Akiyama4,and Hiroshi Matsuzaki5
ABSTRACTThis paper presents damage of road and railway bridges during the Great East
Japan earthquake and tsunami on March 11, 2011 based on a JSCE damage
investigation. Ground motion induced damage and tsunami induced damage of bothroad and railways bridges are presented.
INTRODUCTION
The Great East Japan earthquake (Off Pacific Coast of Tohoku Region, Japan
earthquake) with moment magnitude of 9.0 occurred at 14:46 (local time) on March 11,2011 along the Japan Trough in the Pacific. It was the sixth largest earthquake everrecorded in the world. The fault zone extended 450 km and 200 km in the north-south
and west-east directions, respectively. Extensive damage occurred in the wide region in
the east part of Japan.
The authors were dispatched by Japan Society of Civil Engineers for fielddamage investigation to bridges in Miyagi-ken and Iwate-ken between March 29-April
3, 2011. In addition to the first investigation, damage investigations were conducted
several times. Since 1978 Miyagi-ken-oki earthquake and 2003 Sanriku-Minami
earthquake affected this region, an emphasis was placed in the damage investigation tocompare damage among 2011 Great East Japan earthquake and two previous
earthquakes. This paper presents ground motion induced damage and tsunami induced
damage of bridges during 2011 Great East Japan earthquake.
GROUND MOTIONS AND TSUNAMI
A number of strong motion accelerations were recorded by the National
Institute of Earth Science and Disaster Prevention and Japan Meteorological Agency.
Fig. 1 shows measured accelerations along the Pacific coast. Ground accelerations
continued over 300s, and had at least two groups reflecting the fault rupture process.
The highest peak ground acceleration of 27.0 m/s2 was recorded at Tsukidate. However
the high acceleration was resulted from a single pulse with high frequency components,
and the response acceleration at 1.0 s was only 5.1 m/s2. Damage of buildings and other1

Professor, Tokyo Institute of Technology, Tokyo, Japan
Professor, Kyusyu Institute of Technology, Kita-Kyusyu, Japan
3
Associate Professor, Disaster Prevention Research Institute, Kyoto University, Uji,
Japan
4
Professor, Waseda University, Tokyo, Japan
5
Assistant Professor, Tokyo Institute of Technology, Tokyo, Japan2

䋭23䋭

Photo 1 Shear failure of a column due totermination of longitudinal bars with insufficientdevelopment (Fuji Bridge) (courtesy of Dr.
Hoshikuma, J., PWRI)

Photo 3 Damage of pin and roller bearings

Photo 2 Yuriage Bridge

Photo 4 Damage of PC girders near the support

GROUND MOTION INDUCED DAMAGE OF ROAD BRIDGES
Damage of bridges which were not yet retrofitted
Ground motion induced damage of road bridges was generally less significant.
However extensive damage occurred at the bridges which were designed according to
the pre-1990 design codes (JRA 1990) and were not yet retrofitted in accordance withthe post-1990 design codes. For example, Photo 1 shows shear failure of a reinforced
concrete column resulted from insufficient development at cut-off of longitudinal bars.
This mode of damage occurred extensively in the 1995 Kobe, Japan earthquake[Kawashima and Unjoh 1997]. Extensive investigation was directed to clarify the
failure mechanism of such columns [for example, Kawashima, Unjoh and Hoshikuma
2005], including a project consisting of series of large scale shake table experiments
using E-Defense [Kawashima et al 2009]. Over 30,000 columns were so far retrofitted
since 1995 Kobe earthquake. Consequently, during this earthquake, damage due to thismechanism was not predominant in the bridges which were retrofitted, but damage
occurred at the bridges which were not yet appropriately retrofitted.
Yuriage Bridge as shown in Photo 2 suffered extensive damage at reinforced
concrete hollow and solid columns, supports of prestressed concrete girders, and steel
pin and roller bearings during 1978 Miyagi-ken-oki earthquake. Since the damaged

䋭26䋭

columns were repaired and strengthened by reinforced concrete jacketing, they did not

Photo 6 Damage of a pier during 1978
Miyagi-ken-oki earthquake (Sendai Bridge)

Photo 5 Buckled truss braces

Photo 7 Retrofitted pier of Sendai Bridge which did not suffer damage during 2011 Great East Japan earthquake

suffer damage this time. However pin and roller bearings suffered damage again in thesimilar way as shown in Photo 3. It is obvious that pin and roller bearings arevulnerable to seismic action, because the stress builds up until failure by allowing
virtually no relative displacement at pin bearings and relative displacement
accommodated in roller bearings is insufficient to realistic relative displacement
developed under a strong excitation.
Furthermore, the same supports of prestressed concrete girder which suffereddamage in 1978 Miyagi-ken-oki earthquake suffered again as shown in Photo 4. Taking
account of likely concentration of seismic force at this region and the importance ofanchorage of PC cables, more rigorous repair had have to be conducted after 1978
Miyagi-ken-oki earthquake.
Tennoh Bridge built in 1959 suffered extensive damage during 1978
Miyagi-ken-oki earthquake. This bridge suffered extensive damage again during thisearthquake at the same members; truss braces and pin and roller bearings as shown in
Photo 5.
On the other hand, damage of bridges which were already retrofitted sufferedvirtually no damage. For example, Sendai Bridge which is a symbolic bridge in Sendai
suffered extensive damage at columns and bearings as shown in Photo 6 during 1978
Miyagi-ken-oki earthquake. However this bridge suffered no damage during thisearthquake, because columns were retrofitted as shown in Photo 7 and steel bearings
were replaced with elastomeric bearings.

䋭27䋭

Photo 8 Shin-Tenno Bridge

Photo 9 An end of deck supported by elastomericbearings and unseating prevention devices

Photo 10 Piers where elastomeric bearings ruptured(Sendai-Tobu viaduct)

Photo 11 Transverse offset of a girder due torupture of elastomeric bearings

Photo 12 One of elastomeric bearings ruptured

New bridges constructed by post-1990 codes
Since 1990, the seismic design code was extensively upgraded [JRA 1990].
Before 1990, only elastic static and dynamic analysis was used assuming unrealisticallysmall seismic design force. However after 1990, inelastic static and dynamic analyses
based on Type I/Level 2 design ground motions (middle-field ground motions by M8
events) and an evaluation method of inertia forces considering multi-span continuous
effect were introduced, and introduction of those provisions much enhanced theductility capacity of columns and the seismic performance of bridges. Furthermore in
the 1996 code [JRA 1996, Kawashima 2000] which was revised taking account of

䋭28䋭

damage experience of 1995 Kobe earthquake, the Type II/ Level 2 design groundmotions (near-field ground motions by M7 events), seismic isolation and use ofelastomeric bearings were incorporated. Furthermore, strength of unseating prevention
devices was enhanced.
As a consequence of the upgrading of seismic measures, damage of bridges
which were built or were retrofitted in accordance with the post-1990 design codes
suffered essentially no damage during this earthquake.
For example, Photo 8 shows Shin-Tenno Bridge which was constructed in 2002
suffered no damage. This bridge was located 150 m upstream of Tenno Bridge whichsuffered damage during 1978 Miyagi-ken-oki earthquake and suffered damage again atalmost the same components. Photo 9 shows an end of girder at the left bank where it
was supported by elastomeric bearings. New cable restrainers which satisfy therequirements by the post-1990 design code are set. No damage occurred in this bridge.
Elastomeric bearings including lead rubber bearings and high damping rubber
bearings performed much better than vulnerable steel bearings. However it should benoticed that elastomeric bearings ruptured in several bridges. For example, at
Sendai-Tobu viaduct as shown in Photo 10, several elastomeric bearings ruptured suchthat the deck offset in the transverse direction and settled aside the ruptured bearings asshown in Photo11. Rubber layers detached from steel plates as well as rupture insiderubber layers as shown in Photo 12. Since extensive number of elastomeric bearings
including high damping rubber bearings and lead rubber bearings are used, the damage
should be critically investigated. It is pointed out that one of the possible reasons for thedamage is that the interaction between adjacent bridges with different natural periods
was not properly considered in design of elastomeric bearings. Since an expansion joint
constrained relative displacement between adjacent decks in the transverse direction, it
is likely that larger displacement demand of an adjacent deck is imposed to theelastomeric bearings which were designed based on smaller displacement demand[Quan and Kawashima 2009].
GROUND MOTION INDUCED DAMAGE OF SHINKANSEN VIADUCTS
Seismic retrofit program of Shinkansen
Tohoku Shinkansen started the service in 1982 between Omiya and Morioka
Stations. Since Shinkansen viaducts were designed prior to the occurrence of 1978
Miyagi-ken-oki earthquake, they have smaller amount of shear reinforcements thanrequired by the current code. Some viaducts of Tohoku Shinkansen between Morioka
and Mizusawa-Esashi stations in Iwate-ken were extensively damaged during 2003
Sanriku-Minami earthquake [JSCE 2004].
It should be noted that all viaducts which suffered damage during 2011 Great
East Japan earthquake had not yet been retrofitted. Most viaducts in Iwate-ken were
single story RC moment resisting frame with a Gerber girder at both sides. Damage
concentrated at the side columns during 2003 Sanriku-Minami earthquake as shown in
Photo 13. Since the side columns were shorter than the center column, a parameter
defined as a ratio of the shear capacity to the flexural capacity was smaller in the sidecolumns than the center columns, which led shear failure in the side columns.

䋭29䋭

After 2004 Niigata-ken Chuetsu earthquake [Huyck et al 2006], the first seismic
retrofit program was initiated for Shinkan-sen viaducts including Tohoku Shinkansen.

Photo 13 Single story RC moment resisting framepier with a Gerber girder at both sides

Photo 14 Shear failure of RC columns at Odaki
viaducts after 2003 Sanriku-Minami earthquake
Response Acceleration (m/s2)

20
15

2011 Great East Japan
earthquake
2003 Sanriku-Minami
earthquake

10
5
0
0

1
2
3
4
Natural Period (s)Photo 15 No. 3 Odaki viaducts which were retrofitted Fig. 5 Response accelerations in 2003
after 2003 Sanriku-Minami earthquake performed Sanriku-Minami earthquake and 2011 Great
East Japan earthquake
well during 2011 Great East Japan earthquake

Objectives of the program were to enhance the seismic performance of the columns
with insufficient shear capacity. After retrofitting 12,500 columns, the program wascompleted by 2007. In 2009, the second retrofit program for enhancing columns
flexural capacity was initiated.
No. 3 Odaki viaducts
No. 3 Odaki viaducts in Iwate-ken failed in shear as shown in Photo 14 during
2003 Sanriku-Minami earthquake. The viaducts were retrofitted so that they had
sufficient shear capacity under the first retrofit program after 2004 Niigata-ken Chuetsu
earthquake. The retrofitted columns of the viaducts performed well with almost no
damage during 2011 Great East Japan earthquake as shown in Photo 15.
Fig. 5 compares the 5% damping response accelerations between 2003
Sanriku-Minami earthquake and 2011 Great East Japan earthquake. The records were
measured at approximately 3km from No. 3 Odaki viaducts. Since the fundamental
natural period of a single story RC rigid frame ranges between 0.4s to 0.6s, it isreasonable to considered that the response acceleration of No. 3 Odaki viaducts was䋭30䋭nearly the same between 2003 Sanriku-Minami earthquake and 2011 Great East Japan
earthquake. It is considered that the seismic retrofit for No. 3 Odaki viaducts was

Photo 16 Damage of R7 column, No. 1 Nakasone viaduct during 2011 Great East Japan earthquake(a) Left, R7-1(b) Right, R7-1(c) Left, R7-2(d) Right, R7-2

Photo 17 Close view of damage of R7-1 and R7-2 columns

effective for preventing significant damage during this earthquake.
No. 1 Nakasone viaducts
No. 1 Nakasone Viaducts, constructed in 1978, is located between Kitakamiand Shin-Hanamaki Stations. They had the similar structural shape with the No. 3
Odaki viaducts. No columns were retrofitted during the first seismic retrofit program
since it was evaluated that the parameter was not small enough.
During 2011 Great East Japan earthquake, side columns suffered extensive
damage as shown in Photo 16. All columns lost even the bearing capacity for verticalload. It was fortunate enough not to totally collapse because the viaduct was supported
by eight columns. Side columns in other viaducts also suffered extensive damage.
Shear failure occurred at the upper part of columns as shown in Photo 17. The damage
was so extensive that original shear cracks could not be identified because of crash andspill out of the core concrete.
In the JR seismic evaluation, it was evaluated that shear failure occurred if theparameter was smaller than 0.9. From the fact that the columns in No. 1 Nakasone
viaducts failed in shear although was not as small as 0.9, it is recommended to revise
the criteria of failure mode in the future seismic retrofit program.

䋭31䋭detached from the pier which must have happened if the decks were simply washedaway laterally. It is likely that the decks were uplifted by tsunami buoyancy force andthen they were washed away. Steel plate bearings used in this bridge was very simple
as shown in Photo 22 such that both uplift and lateral force capacities were limited.

䋭34䋭

force which applied from the deck. It is likely that due to tsunami force the deck
uplifted at the sea side first being supported only at the land side, which resulted inlarger tsunami force. Thus the side stopper at the land side collapsed due to excessive
concentration of tsunami force.
Fig. 8 shows possible mechanism of damage of bridges due to tsunami. As
mentioned earlier, overturning of foundation due to scouring did not occur in roadbridges. It is likely that damage of decks in Photos 20 and 21 occurred due tomechanism shown in Fig. 8 (b) and (c), respectively.
In spite of the extensive damage of superstructures, none of piers suffereddamage due to tsunami at Utatsu bridge.
CONCLUSIONS
Damage of road and railway bridges during 2011 Great East Japan earthquake
was presented. Although more through collection of damage information as well as
careful analyses is required, the following conclusions may be tentatively deduced
based on the findings presented herein:
1)

2)

3)

Ground motion induced damage of bridges which were built in accordance withthe post-1990 design code was minor. Thus the effect of enhancing the shear
and flexural capacity as well as ductility capacity, extensive implementation ofelastomeric bearings and strengthening of unseating prevention devises wereeffective for mitigating damage during this earthquake. However, effectivenessof those measures against much stronger near-field ground motions has to becarefully investigated since the ground motion induced by 2011 Great East
Japan earthquake was smaller than anticipated target ground motions.
On the other hand, ground motion induced damage of bridges which were builtin accordance with old code (approximately pre-1990) or which were not yet
retrofitted was still extensive. This was in particular true for railway bridges.
Appropriate seismic retrofit is required in the near future.
Tsunami induced damage was extensive to bridges along the Pacific coast. It
seems that decks were uplifted and washed away upstream. Tsunami force
effect has to be studied more so that it can be considered in design for bridges
along the coast.
ACKNOWLEDGEMENTS

The authors express their sincere appreciation for a number of organizations
and persons for their kind support and cooperation for the field damage investigation.
A joint investigation was conducted as a part of UJNR Panel on Wind and Seismic
Effect activity. Special appreciation is extended to Dr. Yen, F., FHWA and Professor
Buckle, I., University of Nevada, Reno, and members of National Institute of Land andInfrastructure and Public Works Research Institute including Drs. Tamura, K., Unjoh,S., and Hoshikuma, J.

䋭36䋭

REFERENCES
1)

2)

3)

4)
5)

6)
7)

8)

9)

10)

11)

Huyck. C. K., Matsuoka, M., Takahashi, Y., and Vu, T. T.: Reconnaissancetechnologies used after the 2004 Niigata-ken Chuetsu, Japan, Earthquake,
Earthquake Spectra, Vol.22, No. S1, pp. S133-S146, 2006.
Iemura, H., Takahashi, Y., Pradono, M. H., Sukamdo, P., and Kurniawan, R.:
Earthquake and tsunami questionnaires in Banda Aceh and surrounding areas,
Disaster Prevetion and Management, Vol.15, No.1, pp.21-30, 2006.
Ishibashi, T, Ikeda, Y., Kanno, T. and Okamura, H.: The study on earthquake
resistance on design and damage of reinforced concrete viaduct, Journal of
JSCE, No.563/I-39, pp.95-103, 1997.
Japan Road Association: Design specifications of highway bridges, Part V
Seismic design, 1990.
Japan Society of Civil Engineers: Report on reconnaissance damage
investigation to civil engineering facilities resulted from 2011 Great East Japan
earthquake, 2011.
Japan Society of civil engineers: Damage analysis of concrete structures due tothe earthquakes in 2003, Concrete Library, 114, 2004.
Kawashima, K., Unjoh, S. and Hoshikuma, J.: A seismic evaluation method of
RC bridge piers with inadequate anchoring length at termination of main
reinforcements, Journal of JSCE, No. 525/I-33, pp. 83-95, 1995.
Kawashima, K., and Unjoh, S.: The damage of highway bridges in the 1995
Hyogo-ken-nanbu earthquake and its impact of Japanese seismic design,
Journal of Earthquake Engineering, Vol. 1, No. 3, pp. 505-541, 1997.
Kawashima, K.: Seismic design and retrofit of bridges, Key note presentation,
12th World Conference on Earthquake Engineering, pp. 1-20, Paper No. 1818,
Auckland, New Zealand, 2000.
Kawashima, K, Sasaki, T., Kajiwara, K., Ukon, H., Unjoh, S., Sakai, J., Kosa,
K., Takahashi, Y., Yabe, M. and Matsuzaki, H.: Shake table experiment on RC
bridge columns using E-Defense, Proc. 41st Panel on Wind and Seismic Effects,
UJNR, Public Works Research Institute, Tsukuba Science City, Japan, 2009.
Quan, G. and Kawashima, K.: Effect of finger expansion joints on the seismic
response of bridges, Journal of JSCE, A, Vol. 65, No. 1, pp. 243-254, 2009.

䋭37䋭

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䋭38䋭

BRIDGE PERFORMANCE DURING THE GREAT EAST JAPAN
EARTHQUAKE OF MARCH 11, 2011
M. Lee Marsh, Ian G. Buckle, W. Phillip Yen, Shideh Dasti, David Frost, Eric
Monzon (See Acknowledgements for author affiliations)
Abstract
On March 11, 2011 the Great East Japan Earthquake struck the northeast coastal
areas of Japan. Numerous bridges were damaged by either ground shaking or tsunami. A
reconnaissance team from the US visited 11 bridge sites in June, and this report summarizes
the observations and preliminary conclusions of several of those bridges. Staff from the
PWRI hosted the reconnaissance effort and provided valuable technical data, excellent
logistics, and useful insight into behavior and bridge practice throughout the reconnaissanceeffort.
Introduction
About 200 highway bridges and numerous rail bridges were damaged during theGreat East Japan Earthquake of March 11, 2011, including span unseating, foundation scour,ruptured bearings, column shear failures and approach fill settlements. The causes of thisdamage can be broadly classified in two categories: ground shaking including ground failure(liquefaction), and tsunami inundation. Of these, the tsunami was responsible for about onehalf of the number of damaged bridges.
A joint EERI/FHWA/GEER reconnaissance team visited the affected area from June
2 to June 6, 2011 and investigated 11 bridges: 2 had extensive bearing failures, 2 had column
failures, 2 had combined bearing and column failures and 4 suffered tsunami-related damage(unseated spans, scour, loss of approach fill).
The location and names often of the eleven bridges visited by
the reconnaissance team are shownin Figure 1. The eleventh bridge
was the Arakawa Wangan Bridge
across the Arakawa River in Tokyo.
The performance of five of theeleven bridges is summarized in thisreport. A detailed description of alleleven bridges is given in FHWA,
2010.
Bridge Damage Due to GroundShaking
In general the amount ofdamage due to ground shaking was
remarkably light considering peakground accelerations in somelocations exceeded 1.0 g, with

FIGURE 1 – BRIDGES INVESTIGATED BY EERI\FHWA\
GEER RECONNAISANCE TEAM (Graphic: L. Marsh)

䋭39䋭

short-period spectral accelerations in excess of 5g. The most likely explanation is that, most,
if not all of the bridges on the national highway system had been seismically retrofitted overthe last 10-15years (in response to the widespread damage to bridges in the 1995 HyogokenNanbu (Kobe) Earthquake). Bridges damaged in this earthquake by ground shaking weregenerally older structures owned by city and local governments, where retrofit programs havenot been as active due to a lack of funding. With one exception, new bridges performed very
well regardless of ownership most probably due to the adoption of conservative capacity
design principles in the JRA Design Specifications in the 1990s. The one exception was thefailure of elastomeric bearings in a section of the Sendai-Tobu Viaduct as described below.
Three bridges are described in this section.
Sendai-Tobu Viaduct
The damage to this 4.4 km long, multi-span viaduct in north Sendai was largely
confined to a 10-span section between Piers 52 and 62. Built in 2000, this section of the
viaduct was being widened at the time of the earthquake. New on- and off-ramps were underconstruction between Piers 54 and 56 to connect Route 10 carried by the viaduct to Route 141
below. Span lengths and type between Piers 51 and 58 are shown in Table 1.
TABLE 1. SPAN DETAILS AND DISTRIBUTION OF FAILED STEEL STOPPERS AND
ELASTOMERIC BEARINGS IN SENDAI-TOBU VIADUCT.

Span
No.

Span
(m)

Span
Type

52

39.0

8 I-girders

53

54

55

56

71.0

4 box
girders

72.0

4 box
girders

72.0

5 box
girders

71.0

3 box
girders57

39.0

8 I-girders

58

39.0

8 I-girders

59
TOTALS

39.0

Pier
No.

Pier
Type

51

1-col.

No.
failed(damaged)2
stoppers inmain event
3/11/2011052

11-col.

0/6

0/8

0

0

53

1-col..

5

0

0

0

54

2-col.

4(2)

1

0

0

55

2-col.

(6)

0

0

0

561

2-col.

6/6

(1)/8

0

0

57

1-col.

1(1)

0

3

0

581

2-col.

4(2)/5

1/0

0

(2)/0

59

2-col.

(3)

0

0

0

37(14)

18(1)

3

(2)

No.
ruptured(damaged)2
bearings inmain event
3/11/2011
0No.
failed(damaged)2
stoppers inaftershock
3/11/2011
0No.
ruptured(damaged)2
bearings inaftershock
4/7/2011
0

8 I-girders

Note 1. There are two lines of stoppers and bearings on Piers 52, 56, and 58.
2. Numbers in parentheses are numbers of stoppers and bearings damaged but not ruptured.

䋭40䋭

The superstructure
comprises eight steel plate girders(I-girders) between Piers 50 and
52, three, four, or five steel box
girders between Piers 52 and 56,
and eight steel plate girders
between Piers 56 and 63.
Elastomeric bearings are used
exclusively with external stoppers
to restrain transverse movement at
almost every pier. Piers 54, 55, 56,
58 , 59 and 60 had recently been
converted from single steel box
columns to two-column steel box
frames to accommodate the new
on- and off-ramps (Figure 2). The
remaining piers (51, 52, 53 and57) are single-column steel boxes(Figure 3).
The bridge sufferedmoderate-to-major damage
during the earthquake but nospan collapsed. This damage
included the failure of 40 steel
stoppers and 18 elastomericbearings. Another 14 stoppersand 3 bearings were heavily
damaged. In addition, girder
stiffeners, gusset plates, andcross-frames were buckled orseverely distorted. Locations of
the failed and damaged stoppers
and bearings, due to both the
March 11 main shock and theApril 7 aftershock, are given in
Table 1.

FIGURE 2 – TWO-COLUMN FRAME, PIER 56
SENDAI-TOBU VIADUCT (Photo: E. Monzon)

FIGURE 3 – SINGLE COLUMN, PIER 57
SENDAI-TOBU VIADUCT (Photo: E. Monzon)

The pattern of the bearing damage in Table 1 is particularly interesting. It isconcentrated in regions of the viaduct where there is a significant change in lateral stiffness –from single-column hammerhead piers at Pier 57 to two-column frames at Piers 54, 55 and56, for example. There is also a significant change in the in-plane stiffness of thesuperstructure in this section, from eight I-girders in Spans 52 and 57 to multiple single-cellbox girders in Spans 53 to 56. This section of the viaduct is therefore very stiff (andparticularly Spans 55 and 56) while sections to the north and south are comparatively flexible.
When earthquake loads are applied, the difference in displacements at the two interfaces(Piers 52 and 56) generate high lateral forces in the stoppers at these two piers leading to theirfailure and the transfer of load to the bearings.
Inspection of the damage to the bearings showed that many had ruptured completely
through the elastomer, as if in direct tension. Others showed damage to the internal shims
which had been severely distorted (Figure 4). Typical dimensions of the bearings at Pier 56

䋭41䋭

are 820 x 870 x 508 mm, with 8 x 33
mm layers of elastomer, 7 x 4.5 mm
shims and 2 x 45 mm end plates. The
masonry and sole plate connections
were detailed to transfer both shear andaxial forces (tension and compression)
into the bearings.
It seems possible that thebearings failed due to the combinationof two effects. First the high lateralforces in the steel stoppers at Piers 52
and 56 were probably not evenly
distributed amongst the three effective
stoppers. (Although there are six
stoppers at each pier, only three are
FIGURE 4 – DAMAGED ELASTOMERIC
effective in any one direction at any
BEARING FROM PIER 52, SENDAI-TOBU
point in time.) This uneven distribution
VIADUCT (Photo: E. Monzon)
arises because the gaps between the
stoppers and the sole plates of the
bearings are not exactly the same at each location and one stopper will generally engage
before the others. Overloading of this stopper is very likely, leading to its failure, followed by
the transfer of load to the other stoppers which then fail in turn. Once all the stoppers havefailed the transfer of load to the bearings places them under very high shear strain.
The second effect is the generation of high tensile forces in the bearings at these same
locations due to the difference in pier type. For example Pier 56 is a 2-column frame and Pier
57 is a single column hammerhead pier. Under lateral load the hammerhead rotates about alongitudinal axis twisting the superstructure about the same axis. But the pier cap in the 2column frame at Pier 56 does not rotate in this manner and this frame resists the twisting of
the superstructure. High tensile forces are developed in the bearings as a result.
The simultaneous occurrence of high tension and high shear in the bearings could
have led their failure.
The shim damage seen in Figure 4 most likely occurred when a ruptured bearing
impacted a toppled stopper puncturing the cover rubber layers and distorting the edge of theshim plate. On the other hand, the expected failure mode of an elastomeric bearing is rupture
within a rubber layer and not delamination at the shim plate. The clean surface of this plate in
Figure 4 implies inadequate bond between the elastomer and shim during manufacture thus
reducing the bearing’s capacity for combined tension and shear.
Other damage to the superstructure included buckled cross-frame members, gusset
plates and stiffeners, possibly due to the abrupt change in load path where the transverse
member changes from a partial height diaphragm to a full depth cross-frame. But a more
likely scenario is that this damage is due to the failure of the bearings below the girders
leading to differential ‘settlement’ of the cross-frames and corresponding distortion anddistress.
Yuriage Bridge
The Yuriage Bridge carries Route 10 over the Natori River near the Sendai airport.
The area experienced tsunami run-up, but the wave passed under the bridge and the

䋭42䋭

superstructure was not impacted.
The bridge was built in 1974 andis comprised of ten spans with an
overall length 542 m. The threemain spans, located between Piers
2 and 5, are cast-in-place concrete
box-girders and the center of thesethree spans comprises twobalanced cantilevers meeting at
mid span (Figure 5). The approachspans are prestressed concrete Igirders, simply supported on steel
bearings at the pier caps, founded
on wall piers with caisson-type
foundations.
Pier 1 was damaged
during the 1978 Miyagi-Oki
Earthquake and was repaired witha concrete jacket. No widening orother seismic retrofits have been
made since that time.

FIGURE 5 – MAIN SPANS OF THE YURIAGE
BRIDGE (Photo: E. Monzon)

Complete bearing failuresoccurred at Piers 2 and 5, at thetransition between the approachspans and main spans. These
failures are attributed topermanent pier movement
possibly caused by liquefaction.
Evidence of extensive liquefaction
was found under the approachspans and in some cases 30 cm ofground settlement was observed
FIGURE 6 - EJECTED ROLLER BEARING AND
adjacent to one of the piers under
PERMANENT LONGITUDINAL MOVEMENT AT
the south approach. Since Piers 2
PIER 2, YURIAGE BRIDGE (Photo: E. Monzon)and 5 are close to the river, lateral
spreading may have caused thesepiers to move towards the center of the river channel. Figure 6 shows this movement for Pier
2 to be of the order of 6 cm, which clearly exceeded the capacity of the roller bearing at thislocation. In addition damage to the transverse stopper is evident, indicating the simultaneous
occurrence of significant shaking transverse to the bridge.
Shida Bridge and Levee
The Shida Bridge is a nine-span steel plate-girder bridge carrying Route 32 over the
Naruse River east of Osaki. The bridge was built in 1957 and is comprised of a two-girdersteel superstructure supported on concrete two-column piers. The foundation type isunknown. The bridge is straight, has no skew and only a slight vertical curvature. Typicalpiers are shown in Figure 7. The superstructure is articulated in every other span with drop-inspans, whose in-span seats form inflection points rendering the structure determinate (Figure8). Such a drop-in span is visible in the center span of Figure 7 where rust stains from the䋭43䋭Bridge Damage due to Tsunami Inundation
Twelve bridges on Route 45 were seriously damaged by the tsunami, which had wave
heights from Sendai to Hachinohe ranging from 6.2 to 11.8 m. Damage to two of thesebridges and nearby rail bridges are described in this section.
Koizumi Highway and Rail Bridges
The Koizumi bridge spans the Tsuya River on Route 45 just south of the city of
Kesennuma. The bridge was constructed in 1975, has six 30.1-m spans (total length 182 m),
and is 11.3-m wide. The superstructure comprised four steel plate girders supported by
concrete piers on deep foundations. The bridge is without skew and only a slight verticalcurve. The superstructure segments were continuous over three spans with expansion joints at
the abutments and at the center pier (Pier 3). Piers 2 and 4 had fixed bearings, while Piers 1
and 5 had sliding bearings in the longitudinal direction.
The bridge had been seismically retrofitted using hydraulic dampers at the abutments.
It is not known if similar restrainers or dampers had been installed at the expansion joint overthe center pier. No other retrofitting, such as support length extensions or substructure
strengthening was evident.
All six spans were swept away during the tsunami (Figure 13). Wave heights of theorder of 11.8 m were registered at Ofunato City just north of this site and the tsunami clearly
overtopped this bridge taking all six spans upstream. Based on damage to the levee on thenorth bank of the river, some of these spans were lifted off their piers and swept along the topof the levee on the north bank, then over the levee altogether on the north side, and later backover the levee into the main channel where they came to rest about 400 m upstream from thebridge (Figure 14). Other spans took a different path and came to rest
about 300 m upstream but on the south side of the levee on the south bank of the river. Fourof the five piers are still standing, but the center pier (Pier 3) was overturned and believed to
be under water in the river channel just upstream of the bridge (Figure 13).

FIGURE 13 – REMAINING PIERS OF KOIZUMI BRIDGE. TEMPORARY BRIDGE IS UNDERCONSTRUCTION ON SEAWARD-SIDE OF ORIGINAL BRIDGE (Photo: E. Monzon)

䋭47䋭

It is clear that thelongitudinal dampers installed at
the abutments and the transverse
keys (stoppers) over the piers,
offered little restraint to the lateralloads imposed by tsunami. Oncethese devices failed, the relatively
light weight of the steel I-girders,
together with the buoyancy effectsof air trapped between the girders,enabled the superstructure to beeasily lifted and carried significant
distances upstream. The loss ofPier 3 was probably due to scour
but this could not be confirmed.
Despite the low tide at the time ofthe visit, the foundation was still
underwater.
About 900 m upstream of
the Koizumi Bridge, the JR East
rail line to Kesennuma crosses the
Tsuya River on a multispan,prestressed concrete girderviaduct. Five of these spans were
washed out, but the piers survived(Figure 15). The in-coming
tsunami apparently breached thelevee behind the piers allowing
flow oblique to the channel. The
piers are tilted toward the breach,and the simple span, three-girder
superstructures came to rest on the
opposite side of the levee.

FIGURE 14 – GIRDERS FROM KOIZUMI BRIDGE 400m
UPSTREAM IN TSUYA RIVER CHANNEL (Photo: E.
Monzon)FIGURE 15 – DAMAGED PIERS OF THE JR RAIL
VIADUCT AT THE TSUYA RIVER (Photo: S. Dashti)

Of interest is the damage to the lower portions of the piers. The exposed
reinforcement seen on the left side of the each pier appears to have been pulled outward fromthe center of the column and rupturing the relatively light transverse steel. This type of
behavior is seen in the failure of beams that are unreinforced for shear, where a shear crack
precipitates failure and tearing of the tensile reinforcement from the beam. In the case of the
JR East piers, potential buoyancy of the superstructure due to trapped air and thehydrodynamic forces produced lateral loads on the piers along with eccentric vertical loading.
The piers may have failed in shear above the foundation after plastically deforming under the
combined lateral and vertical effects. Following the loss of shear capacity at the base, thetension reinforcement was torn from the piers.
In this postulated mode of failure the tilting of the pier is due to structural failure andnot to scour and subsequent rotation of the foundation. Inspection of the columns and footings
below the water line is required to confirm this behavior.

䋭48䋭

Nijyu-ichihama Highway and Rail Bridges
The Nijyu-ichihama highway bridge spans a small stream on Route 45 south of
Kesennuma and the Koizumi and Sodeo-gawa bridges. This bridge was built in 1971 and is asingle-span prestressed concrete I-girder bridge supported on tall, cantilever abutments, which
are in turn supported on steel pipe piles. The bridge has no skew, no curve and essentially nograde. The span is 16.64 m and the total width of the original structure is 8.7 m. End
diaphragms engage each of the eleven I-girders comprising the deck and in turn, and wereanchored to the abutment seats with tie-down rods in each bay. These same diaphragms acted
as transverse shear keys restraining
the lower flange of each girder fromlateral movement.
The bridge has been
widened on both sides at some timein the past using precast double-tee
beams spanning between new
abutments each founded on steel
piles with heads at a much higher
elevation than those of the original
structure. The tsunami washed out
FIGURE 16 – LOSS OF BACKFILL ON BOTH
the backfill behind both abutments
APPROACHES TO SINGLE-SPAN NIJYU-ICHIHAMA
and temporary approach spans,
BRIDGE (Photo: L. Marsh)using steel I-girders, were placed toopen the road to traffic. These spansare seen in Figure 16. Temporary steel towers to support these spans may also be seen in thisfigure.
Apart from the loss of the seaward extension, this bridge has performed remarkably
well from a structural point of view. It is essentially intact and the principal reason for closure
was the loss of back fill due to erosion. Despite the buoyancy of trapped air, thesuperstructure was well anchored both vertically and laterally to the abutment seats and was
not dislodged by the tsunami
despite being overtopped. It is ofcourse possible that the erosion of
the abutment backfills and theopening up of two alternative
hydraulic channels took load off
the bridge, but nevertheless theperformance of this bridge underthese circumstances is noteworthy.
About 100 meters
upstream from the Nijyu-ichihama
bridge is the JR East line to
Kennesuma, which runs a distanceof several hundred meters acrossthe valley between tunnels at
either end. This section of rail line
was supported on a long fillembankment, two box culvert
roadway underpasses, and aprestressed concrete, single span

FIGURE 17 – EXPOSED WINGWALLS OF JR RAIL
BRIDGE 100m UPSTREAM OF NIJYU-ICHIHAMABRIDGE DUE TO LOSS OF APPROACH EMBANKMENT(Photo: D. Frost)

䋭49䋭

bridge over the river (Figure 17). The unprotected embankment fill appeared to be a granularmaterial. As the wave overtopped the embankment, it displaced the tracks and significantly
scoured and removed the upper 4 to 5 m of the fill. Apart from the loss of the approach fills,
all the bridges in the valley appeared to be intact.
Preliminary Conclusions
The following conclusions are based on observations made, and data recovered,during this reconnaissance exercise. They are, however, of a speculative nature due to thesmall number of bridges investigated and the absence of detailed field data such as foundationand soil details, bearing and tie-down details, superstructure weights, wave heights, andvelocity profiles at each site. These conclusions are therefore likely to change as additional
data becomes available and further studies are completed.
1. Despite the magnitude of this earthquake, bridge damage outside of the coastal zone
was not heavy. This is believed to be partly due to the distance from the epicentral
region, and partly to the fact that a conservative form of capacity design was
implemented in Japan for new bridges in the 1990s. Furthermore an active retrofit
program was undertaken for older bridges following the Kobe earthquake in 1995,
especially on the national highway system.
2. Aftershocks that follow large magnitude main events can, themselves, be large anddamaging. The damage sustained by some bridges was aggravated in subsequent
aftershocks.
3. Retrofitting is an effective means for minimizing earthquake damage in older bridges.
Most of the observed structural damage occurred in older bridges that had not yet
been retrofitted, or only partly so. It is recommended that strong encouragement begiven to owner agencies to accelerate their retrofit programs.
4. With the exception of several spans in the Sendai area, elastomeric bearings
performed well and considerably better than older-style, steel bearings. The reason
for the poor performance of the Sendai-Tobu bearings needs to be determined quickly
for it has widespread implications on their growing worldwide use as movement andisolation bearings.
5. Damage to several older, un-retrofitted, bridge piers was concentrated in thereinforcement termination zone, and this vulnerability should be considered whenprioritizing bridges for retrofitting. Bridges damaged in this manner are susceptible toadditional damage during aftershocks that will lead to longer repair times and more
restrictive load limits or even closure during repair.
6. Design methods to mitigate tsunami damage from inundation should be developed.
Strategies to keep superstructures in place (such as using integral connections andventing trapped air to reduce buoyancy and equalize hydrostatic pressures on deckslabs) should be explored, along with armoring techniques to prevent undue scour offoundations and approach fills. In addition, the cost of deeper foundations should beweighed against the potential loss of a pier and the need for replacement of one ormore spans.

䋭50䋭

7. Until analytical studies are complete it is not known to what extent the duration ofthis earthquake affected the observed damage, but it is expected to have beensignificant. The effect of duration on structural response should be investigated.
Acknowledgements
The team was hosted in Japan by Task Committee G of the UJNR Panel on Wind and Seismic
Effects. Membership was as follows:
U.S. Members:
Ian Buckle, Professor, University of Nevada Reno, Reno
Shideh Dashti, Professor, University of Colorado at Boulder, BoulderDavid Frost, Professor, Georgia Institute of Technology, Atlanta
Lee Marsh, Senior Project Manager, Berger/ABAM Engineers, Seattle
Eric Monzon, Graduate Research Assistant, University of Nevada Reno, Reno
W. Phillip Yen, Federal Highway Administration, Washington DC (Co-chair)Japan Members:
Taku Hanai, Public Works Research Institute, Tsukuba
Jun-ichi Hoshikuma, Public Works Research Institute, Tsukuba
Kazuhiko Kawashima, Tokyo Institute of Technology, Tokyo
Teturou Kuwabara, Public Works Research Institute, Tsukuba (Co-chair)
Hideaki Nishida, Public Works Research Institute, Tsukuba
Keiichi Tamura, Public Works Research Institute, Tsukuba
Shigeki Unjoh, National Institute for Land and Infrastructure Management
The collegiality of members of the Japan team, the free exchange of technical data, theexcellent field logistics, and many useful discussions about bridge performance are gratefully
acknowledged.
Acknowledgement is also made of generous funding provided by EERI, by FHWA underContract DTFH61-07-C-00031 to the University of Nevada Reno, and by GEER under NSF
Award #CMMI-1138203 to University of California, Davis.
References
EERI, 2011. RAPID Proposal- Performance and Resilience Data Recovery from the Great
East (Tohoku-Oki) Japan Earthquake Considering Bridges Buildings and Government andCommunity Response, Proposal to National Science Foundation, Earthquake Engineering
Research Institute, Oakland CA.
EERI, 2011. Bridge Performance During the Great East Japan Earthquake of March 11,
2011, Reconnaissance Report, to be published, Earthquake Engineering Research Institute,
Oakland CA.
FHWA, 2011. Bridge Performance During the Great East Japan Earthquake of March 11,
2011, Reconnaissance Report, to be published, Federal Highway Administration, Washington
DC.

䋭51䋭

䋭52䋭

27th US-Japan
Bridge Engineering Workshop

Session2

Tsunami and Seismic1
Damage Analysis of Bridges Affected by Tsunami due to Great East Japan Earthquake
By Kenji Kosa
Development of Tsunami Design Criteria for Oregon Coastal Bridges
By Bruce Johnson and Solomon Yim
FHWA Seismic Hazard Mitigation Research Program
By W. Phillip Yen

Page Left Blank

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Conclusions
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䋭66䋭

DEVELOPMENT OF TSUNAMI DESIGN CRITERIA FOR
OREGON COASTAL BRIDGES
Bruce Johnson 1
Based on research and documents by Solomon Yim 2 et al

Abstract
This paper describes Oregon Department of Transportation (ODOT) efforts to develop
guidelines for estimating tsunami forces on bridges. ODOT contracted with Oregon State
University (OSU) to conduct two studies. First we needed a model of the potential tsunami wave
characterized by a height, direction and speed. Second, we needed a model to develop uplift andhorizontal forces on a bridges generated from a wave with the three tsunami wave
characteristics. OSU developed a numerical code to perform modeling of tsunami impact on
bridge superstructures on four bridges located on US Highway 101 in the Siletz Bay area on the
Oregon Coast. The numerical results were incorporated into a mathematical formula to provide a
simplified, approximate method for estimating tsunami forces on bridge superstructures.
Introduction
The Oregon coast is vulnerable to large seismic events from the Cascadia Subduction Zone
(CSZ) which shares common seismic characteristics with those at Sumatra that generated large
tsunamis in the Indian Ocean in December 2004. Studies of tsunami deposits and evidences ofcoastal subsidence indicate that an average of large seismic events in CSZ occurs once every
300-500 years (Goldfinger et al. 2003). The most recent large seismic event in the CSZ occurredin 1700; therefore, there is a relatively high probability that a large seismic event will occur inthe near future that could damage structures along the coastal area in the Pacific Northwest.
The bridges along the Oregon Coast are an important part of the transportation system. Any
major damage to these bridges would result in traffic disruption and impede post-event
emergency response. Since these bridges, mostly built in the 1950-70’s, were not designed toresist large seismic or tsunami loads, they are at the risk of being severely damaged during large
seismic events. However, unlike seismic loads, currently there is no specific design standard for
estimating tsunami forces on bridge superstructures in the US in general and in Oregon inparticular. Therefore, an understanding of tsunami impact on bridge superstructures is of majorinterest to the practicing engineering community. Consequently, the Oregon Department ofTransportation (ODOT) initiated a research program to develop guidelines for estimatingtsunami forces on bridge superstructures in the tsunami run-up zone along the Oregon Coast.
The study first developed numerical models to simulate tsunami impact on bridge
superstructures, and calculate reaction forces due to tsunami loads on four selected bridges on
1
2State Bridge Engineer, Oregon Department of Transportation
Professor, School of Civil and Construction Engineering, Oregon State University

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the Oregon Coast. The four bridges – Schooner Creek Bridge, Drift Creek Bridge, Millport
Slough Bridge, and Siletz River Bridge – are located on Highway 101 in the Siletz bay area. The
study also developed a guideline for estimating tsunami forces on bridge superstructures to be
used as preliminary guidance for design of bridges in the tsunami run-up zone. The developed
guidance is based on existing literature and the time-history results obtained from the numericalmodels calculated in the first part.
OSU had conducted a previous case study of tsunami design criteria on the Spencer Creek
Bridge, on US 101 in Oregon, conducted by Nimmala et al. (2006). The Spencer Creek project
was conducted by developing numerical models of tsunami impact on bridge deck to determine
the time-history forces on the bridge by using LS-DYNA software. The analysis is revisited inthis paper to examine the applicability of the guideline developed in the present work.
Tsunami Flow Simulation
The input tsunami flow fields, water surface elevation and water velocity time-histories, for thesimulation models were obtained from tsunami numerical models developed by Cheung andassociates from the University of Hawaii (Cheung et al. 2010). The nonlinear shallow-watermodel by Yamazaki et al (2009) was utilized to capture hydraulic processes – wave overtopping,hydraulic jump formation, and bore propagation – describing flow conditions at the interested
bridge sites.
The development of a rupture model based on 500-year return period CSZ earthquake scenarios
from the National Seismic Hazard Maps. These rupture boundaries extend approximately 1,100km from Cape Mendocino in northern California to Vancouver Island in British Columbia. The
western boundary of the rupture is specified along the trench at the base of the continental slope.
Additional conditions are provided by Wang et al (2003) to define the eastern rupture boundaries
at the midpoint of the transition zone (MT) and the base of the transition zone (BT). Moreover, a
global analog (GA) of shallow-dipping subduction zones, from Tichelaar and Ruff (1993), isused to define the eastern rupture boundary at 123.8qW at 30 km depth.
The tsunami flow model developed by Cheung included four hours of data simulating a 500-year
Cascadia tsunami event at the Siletz Bay for six different scenarios. The six tsunami scenarios
are based on four rupture configurations at moment magnitude (Mw) 9.0 and two additional
moment magnitude 8.8 and 9.2 events at the rupture based on global analog zone. The first
configuration assumes the rupture occurs within the locked zone (LZ) only. The eastern rupture
occurs at the midpoint of the transition zone (MT) and at the base of the transition zone (TZ).
The fourth rupture configuration is assumed to occur at 30 km depth based on global analog(GA).
A relative weight distribution probability of occurrence for the rupture configurations (0.1, 0.2,
0.2 and 0.5 for LZ, MT, BT and GA, respectively) and moment magnitudes (0.6, 0.2 and 0.2 forMw 9.0, 8.8 and 9.2) are assigned based on the logic tree in the Pacific Northwest seismic source
model in Cheung et al. (2010).

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Currently, there is no specific code of practice to estimate forces on bridge superstructures due totsunami loads. However, there is some relevant literature of wave forces on highway bridge
decks and offshore platforms, and some literature on tsunami forces for other types of structures
such as vertical walls, elevated slabs, and columns of different shapes.
Development of Equations to Estimate Forces from Tsunami Waves
Bea et al. (1999) presented a modification of the American Petroleum Institute (API) guidelines
for estimating wind-induced wave forces on a platform deck of offshore structures by separating
the total wave force into two components, horizontal force and vertical force. The horizontal
force consisted of slamming force, drag force, and inertia force. The slamming force and dragforce depended on the horizontal velocity of the waves while the inertia force depended on theacceleration. The vertical force consisted of a buoyant force and a lifting force, which depends
on the vertical velocity of the waves.
Wave Forces on Bridge Decks
Douglass et al. (2006) presented a method for estimating wave forces on typical U.S. coastal
bridge spans due to wind waves and storm surge to offer a preliminary guidance for design
engineers. The estimated horizontal and vertical forces in that method mainly depend on theelevation of the wave crest. Other than water elevation, the horizontal force is also dependant on
the number of girders supporting the bridge deck. This recommended approach was verified with
post-storm damage on U.S. 90 Bridge across Biloxi bay, Mississippi by Hurricane Katrina.

FH[1  C r ( N  1)]( C h  va  C h  im )J ( ' h ) Ah
FV
( C v  v a  C v  im ) J ( ' h ) A v

where
is a reduction factor for forces distribution on the internal girders; is number ofgirders supporting bridge deck;andare empirical coefficients for slow varying
andare empirical coefficients forhorizontal and vertical force respectively;
horizontal and vertical impact force respectively. The other parameters are generally defined innotation.
Previous Research on Wave Forces
FEMA P646 (2008), guidelines for design of structures for vertical evacuation from tsunamis,summarized the relevant design code, and presented equations for estimating tsunami forces onvertical evacuation structures. It also provided some suggestions on how to combine tsunami
force with other loads such as dead load and live load. Load effects that had to be considered fortsunami forces consisted of hydrostatic, hydrodynamic, impulsive, buoyant and uplift forces. The
hydrostatic force depended on water elevation and would be considered to be zero when water
fills up on two opposite sides. Unlike the wave forces due to storm surge, the hydrodynamic
force due to tsunamis depended on flux momentum (hu2) where h is elevation of water crest andu is horizontal velocity. The impulsive force due to tsunami could be estimated by taking 1.5
times the corresponding hydrodynamic force for conservatism.

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Douglass et al. (2006) developed a method for the Federal Highway Administration (FHWA) toestimate wave forces on highway bridge decks due to storm surge. Their approach wasdeveloped based on laboratory experiments of a scaled bridge deck model in a 3D wave basin.
The resulting predictions were shown to be adequate for estimating the wave force induced bystorm as verified by measured field damages from Hurricane Ivan and Katrina. However, theequations presented in that method depended only on wave crest elevation without considering
the importance of water velocity, which is an important factor in tsunamis.
Numerical Models for Tsunami Impact Loads on Bridges
The models are developed to perform numerical testing of tsunami impact on realistic bridge
superstructures to predict the magnitude of tsunami forces that could occur on specific types ofbridge superstructure. This section presents details development of the numerical models, bridge
descriptions as well as time-history of fluid loads on bridge superstructures under various
tsunami flow fields. Effects of different cross-sectional bridge types and the effect of bridge rails
to fluid loads are discussed followed by cumulative probabilities of tsunami forces andoverturning moments. Furthermore, computational efforts are also summarized and presented inthis section.
Two-dimensional (2-D) numerical models are developed using a finite-element based code. The
provided tsunami flow velocities are assumed to be uniform over depth and resolved in the
direction perpendicular to the longitudinal span of the bridge. The cross-section of the bridge
superstructure normal to the longitudinal span is modeled by assuming simply supported under
external girders.
In general, a simulation model consists of two major material parts: a fluid part and a rigid
structure part. The fluid part is a composition of water and air materials which are demonstrated
by appropriate material type combining with equation of states. For computational efficiency, anapproximating rigid body material was used to represent the bridge part and reaction forces are
determined by replacing four rigid elements at supports by elastic material. The OSU study
focused on quantifying the maximum value of the horizontal force, vertical force, andoverturning moment due to tsunami loads on the selected bridges; thus, the simulation began at atime immediately prior to first water impact the superstructure and terminated after obtaining thepeak values of the time-history of the loads.
The Lagrangian-Eulerian coupling algorithm combined with an Arbitrary Lagrangian-Eulerian(ALE) solver was used in the numerical models as it is the most mature formulation to simulate
the problem involving interaction between fluid with high velocity and rigid structure. The basic
concept of the Lagrangian-Eulerian coupling algorithm is to track the relative displacements ofthe corresponding coupling points defined at the interfaced between the Lagrangian surface(bridge superstructure part) and inside the Eulerian elements (fluid part).
FIGURE 1 shows an example of the numerical model of the Millport Slough Bridge developedin this research. The model consists of three material parts: water, air, and bridge parts. Material

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properties for each part – such as material mass density, pressure cut-off, fluid viscosity,
modulus of elasticity, and Poisson’s ratio – are specified appropriately as they are used in the
ALE differential equation and in calculating of interface stiffness. Even though the numericalmodel is two dimensioned, it could be thought of as a three dimensional rectangular crosssection with unit thickness in z-direction. The cross-section is composed of water and air
material parts with a bridge part inside.

FIGURE 1, MILLPORT SLOUGH EXAMPLE SIMULATION MODEL
Example Bridges and Waves
Numerical models of four selected bridges in the Siletz bay are developed for this tsunami load
estimation study. The first is the Schooner Creek Bridge located close to the open channel of thebay facing directly toward the incoming tsunamis. The reference bridge elevation measured at
the support of the lowest (west-most) bridge girder is approximately 18 feet above mean sea
level (MSL).
The second bridge is the Drift Creek Bridge located southeast of the Schooner Creek in a more
open area. The bridge geometry is similar to that of the Schooner Creek Bridge (deck-girder
section) with a smaller cross-sectional width and less number of girders supporting the bridgedeck. The bridge is designed for a 2% slope with a reference elevation of approximately 14 feetabove MSL. The third bridge is the Millport Slough Bridge located at the south end of the Siletz
Bay on Highway 101. The bridge has a 2% slope crown with a reference elevation of 15 feetabove MSL. Finally, the fourth bridge is the Siletz River Bridge. This bridge, which is a box
section, with a reference elevation of approximately 33 feet above MSL, is at a higher elevationcompared to the other three bridges.
Six different tsunami flow fields are provided for each bridge site (GA Mw 8.8, GA Mw 9.0, GA
Mw 9.2, LZ Mw 9.0, MT Mw 9.0 and TZ Mw 9.0). However, the maximum water surface
elevations generated in some scenarios are lower than the reference bridge elevation, and can beneglected because tsunamis in these scenarios would not induce forces on the superstructures. In
particular, five tsunami scenarios – GA Mw 9.0, GA Mw 9.2, LZ Mw 9.0, MT Mw 9.0 and TZ
Mw 9.0 – are applicable to the Schooner Creek Bridge, and the three of these scenarios – GA

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Mw 9.2, LZ Mw 9.0 and MT Mw 9.0 – are also applicable to the Drift Creek Bridge and theMillport Slough Bridge. On the other hand, the tsunami flow of all six tsunami scenarios were
below the beam elevation of the Siletz River Bridge, so no tsunami loads were modeled for thatbridge.
Example Tsunami Force Time-History
Time-histories of the predicted horizontal and vertical reaction forces due to tsunami loads on thethree affected bridges were calculated from the numerical models. The Schooner Creekhorizontal tsunami forces are shown in FIGURE 2. The forces on the box section (black line)
show a pattern of a short duration high intensity force at the time immediately after water
impacting the bridge followed by fluctuating drag forces similar to those reported by Yeh et al.
(2005). The impact forces on the box section are approximately 1 to 2.5 times the corresponding
drag forces; whereas the maximum impact horizontal forces on the deck-girder section are
sometimes smaller than the corresponding maximum drag force. A comparison of the verticaltsunami force time-histories on both box section and deck-girder for Schooner Creek is shown in
FIGURE 3. The vertical tsunami forces on both sections show similar pattern as they are rapidly
increased at the time water impacts the structure followed by steady forces for a while until thewater subsides. Similar results were obtained for Drift Creek and Millport Slough.

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FIGURE 2, SCHOONER CK HORIZONTAL TSUNAMI FORCE TIME-HISTORIES䋭73䋭

FIGURE 3, SCHOONER CK VERTICAL TSUNAMI FORCE TIME-HISTORIES
To summarize, tsunami forces on the superstructure of the selected bridges are quite difference
given the same tsunami scenario. According to the results discussed above, the Siletz River
Bridge could survive a 500-years Cascadia tsunami event because the designed reference
elevation of the bridge superstructure is sufficiently high to avoid tsunami loads while the otherthree bridges are inundated in some scenarios. The Schooner Creek Bridge and the Drift Creek
Bridges were subjected to large tsunami forces, compared to the forces on the Millport Slough
Bridge, because they are located in an open area close to the inlet channel of the bay facing
directly to the incoming tsunamis while the Millport Slough is located far from the inlet channel.
A regression line relating the maximum horizontal forces and the corresponding maximum flux
momentums is plotted in FIGURE 4. It is reasonable to assume that the maximum horizontal

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force is approximately linearly proportional to the maximum flux momentum as suggested in
FEMA (2008) and PBTE (2010).

FIGURE 4, MAXIMUM FLUX (Horizontal axis, in3/sec2) VERSUS MAXIMUMHORIZONTAL TSUNAMI FORCE (Vertical axis, pounds/in)
According to the numerical results, the magnitude of the tsunami forces on a bridge
superstructure generated from different rupture configurations and moment magnitudes can besignificantly different.
Estimation of Tsunami Forces on Bridge Superstructures
This section presents a development of a guideline for estimating tsunami forces onsuperstructures for preliminary design of bridges in a tsunami run-up zone along the Oregon
Coast. This approach is developed by incorporating the relevant existing literature and thetsunami forces obtained from the numerical models developed in the OSU Study for Oregon
DOT.
The total tsunami force on a bridge superstructure can be considered separately as horizontal andvertical components. The horizontal component acts perpendicularly at the center of gravity of
the longitudinal span of the bridge superstructure while the vertical component acts in upwardand downward directions at the center of gravity of the superstructure normal to the wave
direction.

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FIGURE 5, SIMULATED HORIZONTAL AND VERTICAL TSUNAMI FORCES
Horizontal Forces
The total horizontal forces on the bridge superstructures due to tsunami loads are a combinationof hydrostatic and hydrodynamic pressures. The hydrostatic pressure is induced by gravity, andincreases with water depth. The total force due to hydrostatic pressure is a result of imbalanced
pressure, which could be considered zero when water filled up both side of the structure. The
hydrodynamic pressure is induced by horizontal water velocity which is a significant factor in
the tsunami events. The hydrostatic and hydrodynamic forces are considered linearly
proportional to the water elevation and the flux momentum (hu2), respectively.
The total horizontal wave-induced force on bridge superstructures presented by Douglass et al.
(2006) is estimated by combining the hydrostatic pressure on the seaward external girder and thetotal pressure on the internal girders. The total force on the internal girders can be estimated bymultiplying reduction factor with the corresponding force on the seaward external girder. The
horizontal force due to hydrostatic (Douglass et al. 2006) and hydrodynamic (Yeh 2007)
pressures, therefore, can be formulated as shown below.

Fh(1  C r ( N  1)) C h Fh*

F d 0 .5 C d U b ( ' h u 2 ) m ax
Where Cr = 0.4 reduction coefficient for pressure on internal girders; N = number of girder
supporting bridge deck; Fh* == seawater massmax )A h ; Cd = empirical drag coefficient;2
density; and (hu )max = maximum flux momentum.
The total horizontal force due to tsunami loads consists of hydrostatic force (water elevationdependent term) and hydrodynamic force (flux momentum-dependent term). Even though the
maximum of these forces might not occur exactly at the same time, combining these maximum
forces together is considered reasonable (and conservative) for design purpose. Therefore, the䋭76䋭

maximum horizontal force on bridge superstructure due to tsunamis can be estimated by
combining the equations above as follows:

FH

Fh  Fd(1  C r ( N  1)) Fh*  0.5C d U b ( 'hu 2 ) max

An empirical drag coefficient, Cd, for bridge superstructures were evaluated in this researchbased on the time-history results obtained from the numerical models. A plot between the total
horizontal force and flux momentum can be considered separately into two parts. The first part iswhere the horizontal force increases rapidly with a small change in the flux momentum (fluxmomentum-independent part). The second part is where the horizontal force increases
proportionally to the corresponding flux momentum (flux momentum-dependent part) as shownin FIGURE 6. The empirical coefficient was estimated from the slope of the graph between flux
momentum and the total horizontal force as. Therefore, the drag coefficientis approximately 1.0 for the deck-girder bridge type.

FIGURE 6, TOTAL HORIZONTAL FLUX VERSUS FLUX MOMENTUM
In determination of wave forces due to wind wave and storm surge, it is recommended that the
total horizontal pressure on internal girders could be estimated as 40% of the pressure on theexternal seaward girder. However, horizontal pressure time-history results at the bottom ofbridge girders are determined to evaluate an appropriate reduction coefficient for the distributedpressure on the internal girders under tsunami loads. The results (refer to the figures that showthese results) show that the maximum pressure on the internal girders is approximately 20% to

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50% of the corresponding pressure on the external seaward girder. Therefore, the reduction
coefficient, , for this study was taken as 0.4 until further information is obtained.
A comparison between the estimated maximum horizontal forces and the predicted forces
calculated from the numerical models are shown in FIGURE 7. The straight line in thatgraphError! Reference source not found. represents a perfect fit between estimated force andthe predicted force. It can be observed that the estimated forces could be overestimated orunderestimated in some cases because the recommended empirical coefficients are based on anaverage value of the scattering data as shown above.

FIGURE 7, CCOMPARISON OF NUMERICAL PREDICTION OF HORIZONTAL FORCEVERSUS RECOMMENDED FORMULA ESTIMATE
Vertical Forces
Load effects due to tsunamis that must be considered for estimating vertical force under bridge
girders consist of hydrostatic and hydrodynamic pressure. The hydrostatic pressure is induced bywater elevation as mentioned earlier while the hydrodynamic pressure is induced by horizontal

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and vertical water velocity. The summation of estimated pressures under the bridge
superstructure can be estimated by following equation.
1
1
P J ('h)  Uux2  Uuy2
2
2

However, the hydrodynamic force induced by the vertical component of water velocity isrelatively small compared to the corresponding hydrostatic and hydrodynamic forces due tohorizontal velocity; thus, it can be neglected. Consequently, the maximum vertical force due totsunami loads can be estimated by the simplified equation below.
1
FV {J ('hmax )  Uux2,max }Av
2

It is important to remember that these maximum forces might not occur at exactly the same time.
It is considered appropriately conservative to combine these maximum forces together for design
purpose until model testing is conducted to verify the recommendations developed by the
research.
In general, the provided tsunami flow field data – water velocity and water elevation – is basedon tsunami flow without obstruction (which is a bridge superstructure in this study). The resultsfrom the numerical models suggest that the output water elevation and water velocity of tsunami
waves near the bridge are higher than the input values. FIGURE 8 shows a plot between input
value of water velocity and the output value of water velocity obtained from the numericalmodels. The output water velocities are measured near the bottom of the seaward external girderas pressures at this location represent up to 80% of total pressure under the bridge cross-section.
It can be interpreted that the output water velocity near the bridge superstructure isapproximately 3.5 times the input water velocity, based on scattering data shown in FIGURE 8.
The relationship between these input and output water velocity can be formulated as shown in
the following equation.
u x , m a x # 3 .5 u *x , m a x

where= adjusted horizontal water velocity (output water velocity); andhorizontal water velocity.

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= inputFIGURE 8, COMPARISON OF HORIZONTAL WATER VELOCITY WITH ANDWITHOUT OBSTRUCTIONS (BRIDGE SUPERSTRUCTURE)
A comparison of the estimated maximum vertical force and the predicted maximum verticalforce obtained from the simulations is shown in FIGURE 9. The estimated vertical forces areobserved to be overly conservative for small values and slightly under-estimated for large values.
However, the recommended equation is considered appropriate for estimating vertical force due
to tsunamis until further study.

FIGURE 9, CCOMPARISON OF NUMERICAL PREDICTION OF VERTICAL FORCEVERSUS RECOMMENDED FORMULA ESTIMATE䋭80䋭

The maximum percentage values of pressure distribution time-histories under each girder along
the cross-section of the deck-girder bridges are not evenly distributed along the cross-section.
The model found that a maximum 70 to 100% of total pressure is applied to the external seaward
girder and it rapidly decreases for the internal girders. However, the total vertical force isassumed to interact with the bridge at the centroid of the cross-section for simplification at thistime.
Conclusions
The recommended approach is intended to be used for estimating tsunami forces on bridge
superstructures as preliminary guidance for design. This approach is developed by incorporating
those proposed in literature and the time-history of the tsunami forces on bridge superstructures
calculated from the numerical models developed in the OSU research. Given the uncertainties intsunami flow field and lack of laboratory results on realistic bridge model, an appropriate factorof safety should be added into these equations.
The input parameters required for estimating tsunami forces by the recommended approachconsist of maximum water elevation, horizontal water velocity, maximum flux momentum, andelevation of bridge superstructure. Moreover, tsunami waves usually loosen sediment saturated
with seawater while surging inland increasing the effective fluid density above that of typicalseawater. Thus, FEMA (2008) recommended the fluid density be set equal to 1.2 times typicalfreshwater density for tsunami forces calculation.
The recommended empirical coefficients are given here. The reduction factor for forces oninternal girders, , is given as 0.4 which corresponds to that presented in Douglass et al. (2006)
as the maximum fluid pressure on the internal girders is approximately 20% to 50% of thepressure on the seaward external girder. The drag coefficient
was obtained for bridge
superstructures under tsunami loads in this study.
The recommended approach is developed based on the deck-girder bridge section only. It might
not be appropriate to apply these recommended equations directly to calculate tsunami forces onother types of bridge superstructures.
ODOT considers the research to be a good start in developing design criteria. However the
proposed formulas for estimating forces need to be verified and the coefficient needs to becalibrated to actual experience. Wave tank model testing is one way we are considering to refine
the results of the OSU research.
Acknowledgements
This paper is based on research conducted at Oregon State University under contract with
Oregon DOT. The work was led by principle investigator, Solomon Yim, with contributors,
Sutaporn Boon-Intra, Seshu B. Nimmala, Holly Winston, Mohsen Azadbakht. The tsunami
wave simulation was developed by Kwok Fai Cheung, Department of Ocean and Resources

䋭81䋭

Engineering, University of Hawaii. This paper is a summary of research report to be publishedin late 2011 by Oregon DOT.
References
Bea, R.G., Xu T., Stear J., and and Ramos R. "Wave Forces on Decks of Offshore Platforms."
Journal of Waterway, Port, Coastal and Ocean Engineering, 1999: 136-144.
Cheung, K.F., Wei Y., Yamazaki Y., adn Yim, S.C. Modeling of 500-year Tsunamis for
Probabilistic Design of Coastal Infrastructure in the Pacific Northwest. Coastal Engineering,
2011: In Press.
Douglass, S. L., Q. Chen, and J. M. Olsen. Wave Forces on Bridge Decks. Washington: Federal
Highway Administration, 2006.
FEMA P646. Guidelines for Design of Structures for Vertical Evacuation from Tsunamis.
Federal Emergency Management Agency, Washington DC, 2008
Goldfinger. "Deep-water turbidites as Holocene earthquake proxies: the Cascadia subduction
zone and Northern San Andreas Fault systems." Annals of Geophysics, 2003: 1169-1194.
Nimmala, S. B., Solomon C. Y., K. F. Cheung, and Yong W. Tsunami Design Criteria for
Coastal Infrastructure: A case study for Spencer Creek Bridge, Oregon. Technical Report,
Salem: Oregon Department of Transportation (ODOT), 2006.
PBET, Riggs, R., Chock, G., Robertson, I., and Yim, S.C. Development of Performance-Based
Tsunami Engineering, NEES Presentation, 2010.
Tichelaar, B. W., and L. J. Ruff. "Depth of seismic coupling along subduction zone." Journal ofGeophysical Research, 1993: 15829-15831.
Wang, K., Wells, R., Mazzotti, S., Hyndman, R.D., and Sagiya, T. A revised dislocation modelof interseismic deformation of the Cascadia subduction zone. Journal of Geophysical Research,
2003: 108(B1), 2026, doi: 10.1029/2001JB001227.
Yamazaki, Y., and Cheung K.F. Shelf resonance and impact of near-field tsunami generated by
the 2010 Chile earthquake. Geophysical Research Letters, 38, doi:10.1029/2011GL047508, inpress. 2011
Yeh, H. "Design Tsunami Forces for Onshore Structure." Journal of Disaster Research Vol.2
No.6, 2007: 531-536.
Yeh, H., I. Robertson, and J. Preuss. Development of Design Guidelines for Structures that Serve
as Tsunami Vertical Evacuation Sites. Olympia: Washington State Department of Natural
Resources, 2005.

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FHWA SEISMIC HAZARD MITIGATIONRESEARCH PROGRAM

W. Phillip Yen, PhD, PE1

Abstract
This paper describes the Federal Highway Administration’s (FHWA) seismic
research program to mitigate earthquake loss of highway infrastructures. Since 1992,
FHWA has initiated three major research projects in the seismic hazard mitigation;
they are Seismic Vulnerability Study for Existing and New Highway Constructions,
Seismic Vulnerability of Highway System, and the SAFETEA-LU Seismic Research
Program. Major products of these three research programs were introduced and future
developments under the current studies and recommended future studies due to theeffects of recent large devastated earthquakes were also discussed.

Introduction
The public relies on highways for the safe transport of goods and people
across the country. Because roads serve as critical lifelines in the delivery of basic
daily needs, they need to function even in the face of adverse weather and natural
hazards. From 1993–1996, the United States spent approximately $250 million per
week responding to the impacts of natural disasters, with earthquakes, hurricanes, andfloods being the major causes of monetary losses. At times, earthquakes can top thelist. One of the most costly natural disasters in the United States between the late1980s and late 1990s was California’s Northridge Earthquake of 1994, which resulted
in $20 billion in damages.
The loss of life and extensive property damage inflicted by the 1989 Loma
Prieta and 1994 Northridge earthquakes emphasized the need to minimize seismic
risks to the U.S. highway system. Seismic research projects conducted by the Federal
Highway Administration (FHWA) are developing mitigation approaches to reduce
those risks, including a method for assessing seismic risks and various structuraldesigns and retrofitting measures.
Since 1992, FHWA has initiated a series of comprehensive seismic researchstudies targeting retrofitting, design, and risk analysis issues and the seismic research
has produced a number of nationally applicable seismic retrofitting manuals anddesign and risk analysis tools.

1

Principal Bridge Engineer, Office of Bridge Technology, Federal Highway Administration, 1200 New
Jersey Ave. S.E., Washington, DC 20590 USA, Email address: Wen-huei.Yen@dot.gov

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Early Earthquake Mitigation Research
FHWA initiated its earthquake investigations after the 1964 Prince WilliamSound Earthquake in Alaska. FHWA’s followup focused on how bridge engineers
could learn from the Alaska earthquake in terms of geotechnical issues such as soil
properties.
Following the defective performance of bridges during the San Fernando
Earthquake in 1971, FHWA and the California Department of Transportation(Caltrans) began exhaustive studies of the seismic performance of bridges. FHWA
and Caltrans invested $3 million in basic research to develop national guidelines forbridge seismic design. The study evaluated the criteria used at the time for seismic
design, reviewed findings from seismic research for potential use in a new
specification, updated guidelines for seismic design, and evaluated the impact of thoseguidelines on construction and costs.
In 1979, FHWA and Caltrans completed the guidelines, which the American
Association of State Highway and Transportation Officials (AASHTO) adopted in1983 as its Guide Specification for Seismic Design of Highway Bridges. This
specification became the national standard in 1992, following the Loma Prieta
Earthquake. The design philosophy underlying this specification was to prevent
collapse of any span or part of a span during large earthquakes. In small to moderate
seismic events, the code’s intent was for bridges to resist seismic loads without
significant damage to structural components. Under this code, the design earthquake
has a 475-year return period.
ISTEA and the Seismic Research Program
FHWA’s earthquake research did not end with the adoption of this 1992
standard. The agency renewed its commitment to mitigating effects on highway
structures by establishing a seismic research program, as called for in the Intermodal
Surface Transportation Efficiency Act (ISTEA) of 1991. The National Center for
Earthquake Engineering Research, later renamed the Multidisciplinary Center for
Earthquake Engineering Research (MCEER), conducted this program for FHWA.
Under ISTEA, Congress funded the research with more than $14.5 million between
1991 and 1997, and the program covered all major highway system components(bridges, tunnels, embankments, retaining structures, and pavements). Approximately
65 percent of the Nation’s 600,000 highway bridges were constructed prior to 1971,
with little or no consideration given to seismic resistance. In recognition of thatsituation, the FHWA seismic research program initiated two comprehensive studies. In
the fall of 1992, the program began studying the seismic retrofitting of existing
highways, and in spring 1993 began studying the seismic design of new highways.
The first product of this research, Seismic Retrofitting Manual for Highway
Bridges (FHWA-RD-94-052), appeared in 1995 and summarized lessons learned from
more than 20 years of earthquake engineering research and implementation, andprovided procedures for evaluating and upgrading the seismic resistance of existingbridges. In 1999 the program published Impact Assessment of Seismic Design of䋭84䋭

Highway Structures, which became the major documentation used to develop
recommendations for the seismic design of new bridges. In 2006, FHWA issued thefinal products of this research, Seismic Retrofitting Manual of Highway Structures–Part I (Bridges) (FHWA-HRT-06-032) and Part II (Retaining Structures, Slopes,
Tunnels, Culverts, and Roadways) (FHWA-HRT-05-067). These new seismic design
specifications were performance-based, and the major difference between them andthe 1992 design code was that they had a two-level design criterion. The higher level
was based on a 1,000-year return period, and the lower on a 100-year period.
Seismic Research Under TEA-21
In 1998 FHWA launched a congressionally mandated seismic research
program under the Transportation Equity Act for the 21st Century (TEA-21), fundedby another $12 million, to study seismic vulnerability. In cooperation with MCEER,
the program conducted a series of studies to develop tools for evaluating andassessing the social costs and impacts of earthquakes on the U.S. highway system.
The goal was to reduce the likelihood of damage to existing and future highway
structures caused by moderate to significant seismic events.
••••

The main tasks undertaken within this program were the following:
Development of loss estimation methods for highway systems
Preparation of a manual for the seismic design and retrofitting of long-spanbridgesDevelopment of protective systems and a systems design manual for bridges
Specialized ground motion, foundation, and geotechnical studies

Under TEA-21, FHWA worked with the National Cooperative Highway
Research Program (NCHRP) in 2001 to develop new seismic design specifications.
The project number and name was NCHRP 12-49, Comprehensive Specification for
the Seismic Design of Highway Bridges. AASHTO then reviewed and revised the
new design specifications and adopted them in 2007. The NCHRP 12-49 specification
was developed from the 1999 recommendations and is about the same. The 2007
specification is a one-level design criterion for a 1,000-year return period.
With the TEA-21 seismic research program, FHWA developed a software
package called REDARS: Risks from Earthquake DAmage to Roadway Systems toestimate the loss of highway system capacity due to earthquakes. The tool helps
bridge owners estimate how earthquake damages affect post-earthquake traffic flows
and enables them to consider those effects during pre-earthquake planning andprioritizations, and in post-earthquake responses, such as rescue and management ofdamage investigations. The seismic research program released REDARS in 2006.
REDARS is a multidisciplinary tool for seismic risk analysis of highway
systems nationwide. For any given level of earthquake, REDARS uses state-ofknowledge models to estimate seismic hazards (ground motions, liquefaction, andsurface fault rupture); the resulting damage (extent, type, and location) for eachcomponent in the highway system; and repairs of each component’s damage,including costs, downtimes, and time-dependent traffic (that is, the component’s

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ability to carry traffic as the repairs proceed over time after the earthquake).
REDARS incorporates these traffic states into a highway network link-nodemodel to form a set of system-states that reflect the extent and spatial distribution of
roadway closures at various times after the earthquake. REDARS then applies
network analysis procedures to each system-state in order to estimate how theseclosures affect systemwide travel times and traffic flows. Finally, REDARS estimatescorresponding economic losses and increases in travel times to and from key locations
or along key lifeline routes. Users can apply these steps for single earthquakes with nouncertainties (deterministic analysis) or for multiple earthquakes and in estimates ofseismic hazards and component damage (probabilistic analysis).
Although REDARS adequately replicated the performance of the highway
system in the San Fernando Valley during the Northridge Earthquake, much work stillneeds to be done to enable engineers to use the methodology with confidence. Indeed,
the researchers developed REDARS with the expectation that new and moresophisticated modules will be developed over time to improve its accuracy andexpand its range of application.
Also in 2006, the program published the Seismic Retrofitting Manual for Truss
Bridges, particularly addressing truss bridges that are more than 500 feet (152 meters)
long. The program also released a third report, Isolation Bearing Design/Retrofit
Manual, in 2006.
SAFETEA-LU Seismic Research
In 2005, Congress passed the Safe, Accountable, Flexible, Efficient
Transportation Equity Act: A Legacy for Users (SAFETEA-LU). Under the newlegislation, FHWA oversaw $12.5 million in seismic research to work with the bridge
engineering community and enhance the earthquake resistance of U.S. highway
bridges.
Also, SAFETEA-LU mandated a technology exchange and transfer task,
which FHWA conducted through a series of bridge engineering workshops andconferences held nationally and internationally. The meetings involved exchange oftechnical information and performance of cooperative studies.
The outcomes of the succession of programs held over the past four decades
include greater understanding in three areas: seismic vulnerability of specificlocations, geotechnical hazards, and infrastructure vulnerability. Building on thisincreased body of knowledge, FHWA currently is developing improved seismic
designs for new and retrofitted bridges, plus instrumentation to monitor performance.
(1) Assessing Seismic Vulnerability: Hazard Maps
To design a bridge to resist earthquakes, understanding the seismic vulnerability
or earthquake intensity of the bridge’s location is essential. This vulnerability usuallyis described as seismic hazard. The U.S. Geological Survey (USGS) publishes
National Seismic Hazard Maps that display various probability levels of earthquake

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ground motions across the United States. The seismic provisions of building codes,
insurance rate structures, risk assessments, and other public policy provisions
commonly apply probability levels based on the hazard maps.
A 2003 update of the maps incorporates new findings on earthquake groundshaking, faults, and seismicity (that is, how prone a region is to earthquakes). USGS
derived the new maps for a grid of sites across the United States by calculating
seismic hazard curves that describe the frequency of exceeding a set of groundmotions. Currently, the new seismic design and retrofitting criteria for bridges use a
1,000-year return period for a given level of earthquake, which represents not greaterthan a 7 percent probability of that earthquake occurring during a bridge design life of
75 years. AASHTO and USGS issued the new maps and computer software forobtaining seismic hazards by entering ZIP Codes or longitudes and latitudes.
(2) Assessing Geotechnical Hazards
Another factor in designing and retrofitting highway bridges is the geotechnicalhazards that earthquakes can trigger, such as soil liquefaction and settlement, slope
failure (landslides and rockfalls), surface fault ruptures, and flooding. Assessing
geotechnical hazards is a two-part procedure. First, engineers conduct a quick
screening evaluation, generally using information available from field reconnaissance.
If various criteria are satisfied, they consider the risk to be low and require nofurther evaluations. If a hazard cannot be screened out, they conduct more detailed
and rigorous evaluations, which usually require obtaining additional data to assess thehazard and its consequences.
(3)

Assessing Infrastructure Vulnerability
To assess the seismic vulnerability of the U.S. bridge inventory, researchers
often use an indices method to determine a seismic rating. The method involves
assessing a bridge’s structural vulnerability, the site’s seismic and geotechnicalhazards, the socioeconomic factors affecting the structure’s importance, and otherissues such as bridge redundancy and nonseismic structural issues. Through thismethod, researchers arrive at a final, ordered determination of the retrofitting priority
of individual bridges and, ultimately, for the Nation’s entire infrastructure inventory.
The rating system has two parts. The quantitative part produces a seismic ratingbased on structural vulnerability and site hazard. The qualitative part modifies therank in a subjective way that accounts for importance, network redundancy,
nonseismic deficiencies, remaining useful life, and similar issues to arrive at an
overall priority index.
(4)

Mitigation Design of New Bridges
Based on advanced seismic research and experience with destructive
earthquakes, AASHTO, NCHRP, and FHWA have improved seismic designs for new
bridges. The results include design details that directly affect bridge performance
under increased loadings due to earthquakes and other natural hazards.
The performance of U.S. highway bridges in recent large earthquakes has shown

䋭87䋭

that the current state of the art has saved many bridges from collapse caused byunseating of the superstructure or shear failure of the columns,”
The fundamental design objective of current seismic specifications in small tomoderate events is to resist seismic loads within an elastic range without significant
damage to structural components. The objective in large earthquakes is that no span,or part of a span, should collapse. The specifications consider limited damage to be
acceptable in these circumstances, provided it is confined to flexural hinging (that is,a hinge that allows an angle to be adjusted while it remains in place) in pier columns.
Further, damage above ground is preferable so that it is visible in sections of thebridge that are accessible for inspection and repair.
Under current specifications, the seismic performance objective is no collapse
based on a one-level rather than a two-level design approach. The current single-level
design criterion is based on a 1,000-year return period event with not greater than a 5
percent probability of occurring during a bridge’s 50-year design life. As an
operational objective, a bridge’s designers may use a higher, two-level performancelevel, but only with authorization from the bridge’s owners. Current specifications,however, do not provide guidance beyond the one-level approach.
(5)

Seismic Retrofitting of Existing Bridges

Retrofitting is the most common method of mitigating risks; in some cases,
however, the cost might be so prohibitive that abandoning the bridge (total or partial
closure with restricted access) or replacing it altogether with a new structure may befavored. Alternatively, doing nothing and accepting the consequences of damage is apossible option. The decision to retrofit, abandon, replace, or do nothing requires
careful evaluation of the structure’s importance and degree of vulnerability. Limited
resources generally require that deficient bridges be prioritized, with important
bridges in high-risk areas being retrofitted first.
Bridges constructed prior to 1971 in particular need to be retrofitted, based onseismicity and structural types. Toward this end, FHWA issued several publications,
including Seismic Retrofitting Guidelines for Highway Bridges (FHWA-IRD-83-007)
in 1983 and Seismic Design and Retrofitting for Highway Bridges (FHWA-IP-87-06)
in 1987. In 1995, FHWA updated these manuals with more current knowledge andpractical technology; FHWA Seismic Retrofitting Manual of Highway Bridges(FHWA-RD-94-052).
Then, as mentioned earlier, FHWA published Seismic Retrofitting Manual ofHighway Structures–Part I and Part II. This two-volume manual contains thefollowing procedures for evaluating and upgrading the seismic resistance of existing
highway bridges:• A screening process to identify and prioritize bridges that need to be evaluated forseismic retrofitting• A methodology for quantitatively evaluating the seismic capacity of a bridge• Retrofitting approaches and techniques for increasing the seismic resistance ofexisting bridges• A methodology for determining the overall effectiveness of alternative retrofitting

䋭88䋭

measures, including cost and ease of installation
The manual does not prescribe rigid requirements as to when and how bridges
are to be retrofitted. The decision to retrofit depends on a number of factors, several ofwhich are outside the engineering realm. These other factors include, but are notlimited to, the availability of funding and a number of political, social, and economic
issues. A bridge may be exempt from retrofitting if it is located in a seismic zone withvery little ground motion or has limited remaining useful life. Temporary bridges andthose closed to traffic also may be exempt.

Future Developments
The recent huge earthquakes in Japan, Chile, and China have challenged
earthquake engineering communities around the world. The intensity of peak groundaccelerations and long duration of shaking in large earthquakes create greaterdifficulties for designing and retrofitting highway bridges. FHWA’s seismic research
program is exchanging technical information and collaborating on research withseismically active States in the United States and with other countries, including
Chile, China, Italy, Japan, Taiwan, and Turkey.
Over the past 15 years, the program has held a series of conferences aroundthe United States and bilateral workshops with other countries to promote newtechnology and exchange technical information. In 2009, the 25th U.S.–Japan Bridge
Engineering Workshop, held in Tsukuba, Japan, marked the silver anniversary of thistechnology exchange and cooperation.
Under SAFETEA-LU, FHWA is working with MCEER, located at The State
University of New York at Buffalo, and the University of Nevada, Reno, to initiate
two major seismic research studies to help highway infrastructure face the challengesof big earthquakes yet to come. These studies focus on innovative protection
techniques and seismic resilience.
(1) Developing Innovative Technologies
The first study’s objective is to improve the seismic resistance of the U.S.
highway system by developing innovative technologies, expanding their applicability,and developing cost-effective methods for implementing design and retrofitting
technologies. As FHWA applies accelerated techniques to construction of new
bridges and maintenance of existing bridges in high seismicity areas, this study is
attempting to develop more advanced design details to accommodate large groundmotions and increase the mobility and safety of the surface transportation system.
(2) Improving Seismic Resilience
Life-safety (no collapse and no loss of human life) is no longer the sole
requirement for success in designing a highway system capable of resisting theimpacts of a major earthquake. The traveling public now expects resilience as well—

䋭89䋭

that is, rapid recovery and minimal impact on the socioeconomic fabric of modern
society.
The need for resilience has led to development of the concept of performancebased seismic design. Performance measures calculated by REDARS include
congestion and delay times. These measures allow system-level performance criteria
to be specified for earthquakes of various sizes, such as maximum permissible traffic
delay times and minimum restoration times. Thus, these measures allow resilience ofa highway system to be defined and measured in quantitative terms, such as the time ittakes to restore the system’s pre-earthquake capacity. Accordingly, local transportation
authorities can develop financial and societal incentives that will improve resilienceand at the same time reduce risk to life and property.
FHWA and others have made substantial progress in this area, particularly
with respect to the performance of individual components of the built environment,such as buildings and bridges. But the real potential for performance-based design
comes when these concepts are applied to systems and subsystems of theinfrastructure, such as transportation networks, subject to both service load conditions
and extreme events.
This project’s objective is to study the resilience of highway systems with aview to improving performance during major earthquakes. By improving current loss
estimation technologies such as REDARS, the FHWA researchers will develop acomprehensive assessment tool to measure highway resilience. They will identify
factors affecting system resilience, such as damage-tolerant bridge structures andnetwork redundancy, and develop design aids for curved bridges and structures innear-fault regions. To the extent practical, they will implement the new methodologies
and technologies in REDARS and conduct outreach to improve seismic safety.
Concluding Remarks
The greatest difficulty in mitigating earthquake hazards is that seismic eventsoccur without any notice and without any way of accurately predicting when they will
occur, nor what their magnitude will be. Earthquakes are devastating, often resulting
in a large number of deaths, injuries, and extensive infrastructure damage. These
losses occur within minutes. Systematic approaches to evaluating earthquake risks,
including indirect losses such as economic impacts, have become an important issue
to the engineering community. Hazard mitigation methods to reduce earthquake losses
require an enormous effort for development and implementation. FHWA is working
closely with AASHTO and NCHRP to mitigate earthquake hazards and reduce losses,
and the efforts to implement all practical measures to enhance the safety and mobilityof the highway infrastructure are in a race against time with earthquakes.
Reference:
FHWA Congressional Seismic Research Studies under the ISTEA, TEA-21,
and SAFETEA-LU Program.

䋭90䋭

27th US-Japan
Bridge Engineering Workshop

Session3

Seismic2
Failure Investigation and Retrofit Strategies of a 22-Span Steel Girder Bridge during the 2010
Chile Earthquake
By Genda Chen, Zuocai Wang, Phillip W. Yen, and Ian Buckle
Earthquake Response of Highway Bridges Subjected to Long Duration Seismic Motions
By Shojiro Kataoka
Experimental Investigation of Influence of Live Load on Seismic Response of a Horizontally
Curved Bridge
By Ian G. Buckle, Hartanto Wibowo, Danielle M. Smith,and David H. Sanders
Discussion Method by Using Dynamic Analysis for Quakeproofing Long-Span Bridges
By Shunji Nagahashi, Takumi Nishikawa, Mutsuji Kawakami, and Tatsumi Ujie

Page Left Blank

䋭92䋭

FAILURE INVESTIGATION AND RETROFIT STRATEGIES OF A 22-SPAN
STEEL GIRDER BRIDGE DURING THE 2010 CHILE EARTHQUAKE
Zuocai Wang1, Genda Chen2, Phillip W. Yen3, and Ian Buckle4

Abstract
This paper presents a case study on the Cardinal Raúl Silva HenĚiquez Bridgethat was significantly damaged during the Chile Earthquake on February 27, 2010.
Field observations and finite element simulations indicated that the bridge failed
mainly because the excessive longitudinal load of 22 continuous steel-girder spans wastransferred from the girders to their bearing masonry plates at two abutments with aweld connection detail, locally bending the girders due to axial load eccentricity.
Parametric studies demonstrated that an effective retrofit strategy can be developed by
reducing the number of continuous spans, modifying the connection detail, andincreasing the capacity of girders with enlarged bearing seats, additional stiffeners forgirders, and thicker flanges and webs.
Introduction
On February 27, 2010, the M8.8 offshore Maule earthquake occurred on athrust fault along the boundary between the Nazca and South American tectonic plates[USGS, 2010]. The Chile earthquake damaged about 200 bridges and led to 20
collapses. One of the significantly damaged bridges is the Cardinal Raúl SilvaHenĚiquez Bridge. Based on the field inspections by Chen et al. [2010], bottom flanges
and webs of the steel girders at abutment supports and the girder-to-abutment weld
connections were severely fractured during the earthquake. This type of damage isindicative of the presence of excessive longitudinal loads in the bridge superstructuresince only one center expansion joint exists in the entire bridge structure. In addition,
the bottom flanges of the girders were welded on masonry steel plates at abutments.
This detail is not representative to the common practice in continuous steel girder
bridge constructions where a continuous bridge section is simply supported on one
fixed bearing and multiple expansion bearings [Saadeghvaziri et al., 2000]. The
specific girder-to-abutment connection detail in the Cardinal Raúl Silva HenĚiquez
Bridge attracted not only axial and shear forces but also a bending moment in thelongitudinal plane of the bridge under earthquake loads.
1

Ph.D. Candidate, Dept. of Civil, Architectural, and Environmental Engineering,
Missouri University of Science and Technology, Rolla, MO, USA
2
Professor, Dept. of Civil, Architectural, and Environmental Engineering, Missouri
University of Science and Technology, Rolla, MO, USA
3
Research Structural Engineer, Turner Fairbank Highway Research Center, Federal
Highway Administration, McLean, VA, USA
4
Dept. of Civil and Environmental Engineering, University of Nevada, Reno, NV,
USA

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Previous research concluded that steel bridge superstructures are susceptible todamage under low or moderated earthquakes and even more fragile than concrete
superstructures if designed improperly [Itani et al., 2004]. For example, bearing failurein the steel girder bridges that were designed without seismic considerations has beencommonly seen during previous earthquake events due to insufficient seat length. The
load path and the capacity of a bridge system and its individual components at end
supports must be evaluated case by case.
The objectives of this study are to investigate the failure mechanism of girders
and end bearings at two abutments of the Cardinal Raúl Silva HenĚiquez Bridge during
the 2010 Chile Earthquake and develop various effective retrofit strategies for themthrough sensitive studies with a finite element model of the bridge.

Bridge Description and Field Observed Damage(1) Configuration of the Bridge

Sa (g)

Built in 2002, Cardenal Raúl Silva HenĚiquez Bridge is a 22-span, steel-girder
structure crossing the Maule River near Constitución in the NE-SW direction. Eachspan length is 41.5 m. The bridge is supported by two seat-type abutments andtwenty-one intermediate bents. As partially shown in Fig. 1, the first five bents fromthe NE abutment are supported on two reinforced concrete (RC) columns and drilledshafts. The next six bents are supported on three RC columns and drilled shafts. The
following eight bents are steel pile bents with three legs (one vertical and two inclined)with horizontal struts and diagonal braces interconnecting the legs in each bent. The
last two bents are supported on three RC columns that rest on footings. The bridge
superstructure is comprised of two continuous 11-span-long segments with three
expansion joints at the two ends and in the middle of the bridge. It is connected to thebridge substructure by elastomeric pads at all piers to allow for longitudinal movement
except for two abutments. At each abutment, three girders were welded to their bearing
masonry plates that are anchored into the abutment.

Period (sec)Fig. 1 Cardenal Raúl Silva HenĚiquez Bridge

䋭94䋭

Fig. 2 Acceleration Response Spectra(2) Acceleration Response Spectra
The 2010 Chile earthquake generated ground motions with long duration andmultiple pulses [Boroschek et al., 2010]. The three-component earthquake groundmotions at the Hospital Station in Curicó near the bridge site were successfullyrecorded but not released to the public as of today. However, their acceleration
response spectra were made available as shown in Fig. 2. It can be seen from Fig. 2 that
the NS and EW acceleration spectra are much larger than the vertical acceleration
spectrum in a period range from 0.2 sec to 0.6 sec and from 0.8 sec to 1.5 sec. On theother hand, the natural periods of the particular steel-girder bridge almost fall in thisrange. In other words, the bridge is more sensitive to the longitudinal and transverse
motions.
(3) Field Observed Damage
During the earthquake, the NE portion of the bridge moved transversely fromwest to east as shown in Fig. 3(a). All steel stoppers were deformed and girders weredisplaced from their elastomeric pads, resulting in the web and flange bending of theexterior girder about its weak axis. At the NE abutment, the webs and bottom flanges
were fractured in all three girders, and both bearing stiffener and web buckled as
illustrated in Fig. 3(b). The cause of this type of damage is indicative of excessive
longitudinal loads in the superstructure that was resisted by the weld bearing
connection at the abutment. At the SW abutment, the welds from the girder bottom
flanges to the masonry plates were fractured [Chen et al., 2010].

(a) Girder offset and cross frame buckling(b) Girder fracture at the north abutment

Fig. 3 Superstructure Damage at the NE Portion

Response Spectrum Analysis of the Bridge System(1) The Grillage Finite Element Model
A typical cross section of the bridge is schematically shown in Fig. 4. The

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bridge has a total of 12 types of cross sections with various flange widths andthicknesses. The flange width of steel girders varies from 0.25 m to 0.62 m, and thethickness changes from 12 mm to 40 mm. The height and thickness of girder webs are
2.06 m and 12 mm, respectively. The bridge has two types of steel diaphragms spaced
every 2 m. It also has two types of piers: concrete and steel pile bents. The diameter ofcircular concrete columns is 1.5 m while the outer diameter of the 12-mm thick thinwalled circular steel pile is 1.0 m. The concrete bridge deck is 10 m wide and its
thickness is 0.25 m.

10 m

4.34 m4.34 mFig. 4 Bridge Cross Section
To understand its complicated seismic behavior, a grillage mesh finite element
model was established to represent the global responses of the entire bridge, and adetailed three-dimensional (3-D) model was created for the local failure analysis of thefractured girder portion. In the global grillage model as shown in Fig. 5, the beam
elements were used to represent the bridge decks and other components (girders, bent
caps, and columns). The longitudinal beam elements represent the centerline of bridge
members passing through the neutral axis of all cross sections. The moment of inertia,
cross sectional area and unit mass of the longitudinal grillage members were
determined from the properties of as-built girders and a portion of bridge decks basedon the effective width of composite beams [Eugene and Damien, 2005].
Longitudinal grillage beamTransverse grillage beam

Fig. 5 Grillage Modeling of the Bridge
All transverse grillage members were placed at the location of diaphragms.
Their properties were directly calculated from the as-built diaphragm frames. The

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equivalent cross area and moment of inertia of each diaphragm frame were calculated
based on the equivalent displacement criterion. The properties of bent caps andcolumns (reinforced concrete or steel tube) were estimated based on their as-built crosssections. The cross sectional areas and moments of inertia of various grillage members
are presented in Table 1. Here, Ix and Iy represent the moments of inertial about thestrong (horizontal) and weak (vertical) axes of a cross section, respectively. The
Young’s Modulus of Elasticity for the longitudinal and transverse grillage beams andsteel tubes is 210 11 N/m2, while that for the concrete bent caps and columns is
2.510 10 N/m2.
Table 1 Section Properties of Grillage Beam Elements
Grillage beams
1
2
3
4
5
6
Longitudina
l
7
8
9
10
11
12
1
Transverse
2
Concrete column
Steel pipe

Area (m2)
0.17
0.18
0.19
0.19
0.19
0.18
0.17
0.19
0.21
0.17
0.19
0.18
0.003
0.007
1.78
0.04

Ix (10-3 m4)
51.4
92
107
119
123
99
84
116
159
78
130
98
0.002
4.67
249
4.55

Iy (10-3 m4)
3.3
3.5
3.7
3.9
4
3.5
3.4
3.8
4.4
3.4
4.2
3.5
0.0002
0.001
249
4.55

To take into account soil-foundation-structure interaction, all columns in eachbent were simply considered to be fixed at certain depth that can be estimated by[Davisson and Robinson, 1965]:

Lf = 1.8

E1 0.2nh(1)

in which and are the modulus of elasticity and the moment of inertia of columns,
respectively, and nh is the coefficient of the horizontal subgrade modulus of soilmaterials. The grillage finite element model of the entire bridge was set up with
SAP2000 as shown in Fig. 6.

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Fig. 6 Finite Element Model of the Bridge based on Grillage Method(2) Response Spectrum Analysis
The complete quadratic combination (CQC) rule was applied to combine theeffects of modal responses [Chopra, 2007]. Therefore, the peak value of a structural
response can be written as:(2)
where N represents the total number of vibration mode of engineering interest; r0 is aandare the peak responses of the ith and nth
peak response of the bridge system;
modes of vibration, respectively; andis the correlation coefficient between the twomodes.
The natural frequencies of the first transverse, first vertical, and first torsionalmodes of vibration are listed in Table 2. The maximum axial force, bending moment,
and shear of the NE end girders where significant damage was observed are presented
in Table 3. It can be seen from Table 3 that the maximum axial force is 8750 kN mainly
due to the longitudinal earthquake ground motion. Due to load transfer eccentricity (2m) at each end of the bridge, the axial force causes a significant bending moment at theend of each girder, which is most likely underestimated in design. This moment isalmost 6 times as large as that of the bending moment directly caused by the earthquake
load.
Table 2 Natural Frequencies of the First Transverse, Longitudinal, Vertical andTorsional Modes
1st Mode
Natural frequency (Hz)Transverse
0.45

Longitudinal
1.21

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Vertical
1.85

Torsional
2.05

Table 3 Maximum Axial Force, Bending Moment, and Shear at the NE End Girders
Axial force (kN) Moment (kN-m)
8750
2970

Shear (kN)240

3D Finite Element Model for the Fracture Analysis of Girders
To better understand the stress concentration around the fracture location ofgirders, the area of crack initiation, and the process of failure, a 3D finite element
model of a small portion of girder (including bearing stiffeners) was established withABAQUS. Considering 0.7 m fillet weld on the bottom flange of each girder at theabutments or 0.7 m bearing seat length, the detailed 3D model is selected to be 1.7 mlong as schematically illustrated in Fig. 7(a). The steel girder and reinforced concrete
deck were modeled by plate and solid elements, respectively, as shown in Fig. 7(b).
The portion of the bottom flange of the girder, welded on the masonry plate at
abutments, was fixed in the fracture analysis of girders. The flange width and thicknessof the girder are 0.28 m and 12 mm, respectively. The web and stiffener thicknesses are
12 mm and 20 mm, respectively. The maximum axial force obtained at the end of
girders from the global grillage model of the bridge system was divided into two
components based on the weight ratio between the bridge deck and steel girders. Eachcomponent was uniformly distributed and applied on its respective deck or girder crosssection.
To understand the crack initiation and failure process, four (4) load cases wereconsidered for elastic-plastic analysis: 130, 300, 440, and 700 kN. The stress
distribution for each case is presented in Fig. 8. It can be observed from Fig. 8 thatdamage likely initiated at the bottom flange of the girder under an axial load of 130 kN.
Cases 2 and 3 in Fig. 8 indicate that the maximum stress will extend into the girder web
in the area of observed damage after the Chili Earthquake. Eventually, the girder failed
at an axial force of 700 kN, which is 8% of the maximum axial load obtained from theresponse spectrum analysis. Based on the stress distributions under various load cases
and field observations, it was verified that the web fracture was indeed caused by theexcessive longitudinal load, which can be estimated to be approximately 12 times of
the actual capacity of the steel girder bearing system.

䋭99䋭

(b) 3D Model(a) Steel GirderFig. 7 Modeled Portion of Steel Girder and ABAQUS 3D Finite Element Model(a) Case 1: Axial force=130 kN(b) Case 2: Axial force=300 kN(c) Case 3: Axial force=440 kN(d) Case 4: Axial force=700 kN

Fig. 8 Stress Distribution under Each of Four Axial Forces

Parametric StudyIn this section, several potential retrofit strategies are investigated either byreducing the longitudinal earthquake load or increasing the seismic capacity to ensure
a smooth transfer of the seismic load from the girder to abutment. To limit the scope of

䋭100䋭

work, it is assumed that both abutments are adequate to transfer the seismic loads fromthe bridge superstructure to ground. Example retrofit strategies for the bearing
connections include:
1. Reducing the longitudinal seismic load by increasing the number of joints inbridge superstructure (both deck and girder) so that bearing elements between
the super- and sub-structure can transfer the seismic load satisfactorily,
2. Changing the bearing connection detail between the bottom flange of girders
and the masonry plate of abutments, and3. Increasing the seismic capacity of the girder-to-abutment system by increasing
the thicknesses of web and bottom flange, the number of stiffeners, and thebearing area at abutments.
More advanced retrofit strategies include the use of base isolators at each bent
and abutment so that the longitudinal load on each bent can be regulated based on its
available capacity for load transfer, and the use of passive energy dissipation systems
so that earthquake energy can be dissipated. In what follows, only the three strategies
that are detailed above are discussed.
(1) Number of Joints

Axial force (KN)

One effective way to reduce the longitudinal seismic force is to increase thenumber of joints at intermediate bents. Towards this endeavor, the number ofcontinuous spans was considered to be 1, 2, 3, 4, 5, and 6. In each case, the longitudinal
force calculated from the response spectrum analysis based on the grillage beam modelis shown in Fig. 9. It can be clearly seen from Fig. 9 that the axial force is significantly
reduced as the number of continuous spans is reduced.

Number of continuous spans

Fig. 9 Axial Force at the End Girder of Bridge with Various Continuous Span Numbers
On the other hand, more expansion joints in bridge deck mean more
maintenance since these are often the areas for water leakage and corrosion developed
over the years. Furthermore, as the number of expansion joints increases, the redundanteffects for extreme loads diminish. Therefore, there must be a tradeoff between seismic
performance and maintenance cost in practical applications or advanced retrofitting
strategies with dampers and isolators can be developed.

䋭101䋭Table 5 Axial Force Capacities of Retrofitted Bearing Connections with Three Added
Stiffeners (kN)
Plate thickness
(mm)
12
25
50

Bearing seat length (m)
0.7
1
1.2
2137
3215
3907
2961
4599
5843
4520
7192
9329

Demand vs. Capacity (kN)In retrofitted design, the seismic capacity of the structure must be larger than
the seismic demand by a certain safety margin. As indicated in Tables 4 and 5 for the
Cardenal Raúl Silva HenĚiquez Bridge, only when the thickness of the web and flange
is increased to 50 mm, three 30 mm thick stiffeners are added at the girder end, and thebearing seat is lengthened to 1.2 m can the retrofitted bearing connection transfer theexcessive longitudinal earthquake load from the girder to abutment without requiring
additional expansion joints over intermediate bents. If the number of continuous spansis reduced, the longitudinal earthquake load can be significantly reduced and thus morebearing connection retrofitting options in Tables 4 and 5 are viable in design. For
instance, Fig. 11 shows the seismic demand versus seismic capacity for various
combinations of reducing span numbers and increasing web/flange thickness andbearing seat length; when the bridge girders are simply supported, the current bridge
design can transfer the earthquake-induced load. For a specified continuous spannumber, the seismic demand (axial load) can be determined from a global bridge
analysis as shown in dash line in Fig. 11. Various retrofitting options as shown in solid,
dotted, and long dash dotted lines in Fig. 11, which can increase the seismic capacity(axial force) above the seismic demand, can be considered as viable designs withadequate performance.

Bearing length=0.7 m

5 spans
4 spans
3 spans

1.0 m

2 spans
1 span

1.2 m

25
40
Web and flange thickness (mm)(a) With no Added Stiffener

䋭103䋭

Demand vs. Capacity (kN)

Bearing length=0.7 m
1.0 m

5 spans
4 spans
3 spans

1.2 m
2 spans
1 span25
40
Web and flange thickness (mm)(b) With Three Added Stiffeners
Fig. 11 Seismic Demand (Load) versus Seismic Capacity (Force) for Various
Continuous Span Numbers and Bearing Connection Retrofit Designs
Fig. 11 also indicates that if the bearing weld connection were changed to a pin
connection, the bending moment due to the welded bottom flanges of girders at
abutments would disappear and the bending moment due to load eccentricity would beapproximately 8750-2970 = 5780 kN. Indeed, re-running the grillage beam model withpin supports at two abutments gave a moment of 6800 kN. Even with pin supports, thebridge must be retrofitted further by increasing seismic capability such as adding three
stiffeners and/or increasing web/flange thicknesses. The main difference between thepin and fixed supports, however, lies in the design of masonry plates and abutments.
With pin supports, the masonry plates are subjected to both axial and shear forces only,
which can be significantly less demanding in comparison with the fixed supports asseen in the current bridge design.

Conclusions
To understand the main causes of the steel girder fracture and weld fracture ofa 22-span Cardenal Raúl Silva HenĚiquez Bridge during the 2010 Chile Earthquake, agrillage beam model of the entire bridge system and a 3D finite element model of the
bearing connection at bridge abutments have been developed. The representative
acceleration response spectra at the Hospital Station in Curicó near the bridge site wereused for the global bridge analysis. Parametric studies were conducted to investigate
the effects of various retrofit designs at bearing connections. Following is a summaryof the preliminary findings from this study:
1. The current bridge design has 11 continuous spans that are all rested onelastomeric bearings except for end bearing connections at abutments. The
elastomeric bearings allow for some longitudinal displacement under thelongitudinal component of ground motions. Therefore, the fixedgirder-to-abutment bearing connection attracted the longitudinal load of allcontinuous spans, causing fracture damage at abutments.
2. Reducing the number of continuous spans can significantly reduce thelongitudinal load applied at the bearing connection system. However, the䋭104䋭

current bearing connection design is still inadequate to transfer the longitudinalload unless all girders are simply supported. Their combinations may benecessary to transfer the loads induced by a mega earthquake such as the 2010
M8.8 Chile Earthquake.
3. Increasing the web and flange thickness of girders, number of stiffeners, andlength of bearing seats at the bearing connection are all effective measures forseismic retrofitting of the bridge.
4. For the multi-span bridge structure, the longitudinal component of the 2010
Chile Earthquake induced ground motions caused more significant damage
than the vertical earthquake motion. Therefore, due considerations must be
taken of the longitudinal ground motion effect in bridge design and a decision
may have to be made to trade off the seismic design load and the seismic
redundancy in lieu of the number of continuous spans.

References
USGS,

Magnitude 8.8 – Offshore Maule, Chile, February 27, 2010,
http://earthquake.usgs.gove/earthquakes/eqinthenews/2010/us2010tfan/,
Earthquake Hazards Program, 2010.
Chen, G.D., Yen, P. W., Buckle, I., Allen, T., Alzamora, D., Ger, J., and Arias, J. G.:2010 Chile Earthquake Implications to the Seismic Design of Bridges,
Proceedings of the 26th US-Japan Bridge Engineering Workshop, pp. 203-216,
2010.
Saadeghvaziri, M. A., Yazdani-Motlagh, A. R., and Rashidi, S.: Effects ofSoil-Structure Interaction on Longitudinal Seismic Response of MSSS Bridges,
Soil Dynamics and Earthquake Engineering, vol. 20, No. 1-4, pp. 231-242,
2000.
Itani, A. M., Bruneau, M., Carden, L., and Buckle, I. G.: Seismic Behavior of Steel
Girder Bridge Superstructures, Journal of Bridge Engineering, Vol. 9, No. 3,
pp. 243-249, 2004.
Boroschek, r., Soto, P., Leon, R., and Comte, D.: Terremoto Centro Sur Chile-27 deFebrero de 2010, Informe Preliminar No 4, Unversidad de Chile-Depto.
Ingenierial Civil, and Depto. De Geofisica, 2010, 29pp.
Eugene J. and Damien L. K.: Bridge Deck Analysis, Taylor & Francis e-Library, 2005,
278pp.
Davisson, M. T., and Robinson, K. E.: Bending and Buckling of Partially Embedded
Piles, Proceedings of 6th International Conference on Soil Mechanics andFoundation Engineering, Canada, Vol. III, pp. 243-246, 1965.
Chopra, A. K.: Dynamics of Structures-Theory and Application to Earthquake
Engineering, Prentice-Hall, Inc., New Jersey, U.S.A., 2007, 876pp.

䋭105䋭

䋭106䋭

EARTHQUAKE RESPONSE OF HIGHWAY BRIDGES SUBJECTED TO
LONG DURATION SEISMIC MOTIONS
Kataoka Shojiro1
Abstract
Strong motion records obtained during the 2003 off Tokachi, Japan, earthquake(Mw8.0) and the 2010 Maule, Chile, earthquake (Mw8.8) are used to investigate
effects of duration of seismic motion on earthquake response of highway bridges. Not
much difference was found between the responses excited by the seismic motions fromthe two earthquakes despite difference of the duration.
Introduction
Current Japanese design specifications (Japan Road Association, 2002) require
highway bridges to be checked if the bridges satisfy target seismic performances
against Level 1 and Level 2 earthquake motions. Level 1 earthquake motion covers
ground motion highly probable to occur during service period of bridges and its target
seismic performance is set to have no damage. Level 2 earthquake motion is definedas ground motion with high intensity with less probability to occur during the service
period of bridges. The target seismic performance against Level 2 earthquake motion
is set to prevent fatal damage for bridges with standard importance and to limit damage
for bridges with high importance.
There are two types of Level 2 earthquake motion, i.e. Type I and Type II
earthquake motions. Type I represents ground motions from large-scale plateboundary earthquakes, while Type II from inland earthquakes and directly strike thebridges. These design earthquake motions are defined as design acceleration response
spectra with damping ratio of 0.05. Time history waveforms are also shown in thedesign specifications as examples for seismic design using dynamic response analyses.
The time history waveforms were produced by spectral fitting using strong motion
records as original waveforms; their acceleration response spectra were adjusted to fit
to the design spectra by means of a spectral fitting technique.
As for Type I earthquake motion, strong motion records obtained during plateboundary earthquakes of which magnitudes ranging from 7.4 to 8.2 were used asoriginal waveforms. Duration of the example waveforms are up to only 55[s] (durationin this paper will be defined in the next section). Ground motion records with long
duration, however, were obtained during the 2003 off Tokachi, Japan, earthquake(Mw8.0) and the 2010 Maule, Chile, earthquake (Mw8.8). Besides, it has been pointed
out that super-giant earthquakes, of which magnitudes are as large as 9, may occur in
Suruga-Nankai trough, south-western Japan, in the near future – though the 2011 off
the Pacific coast of Tohoku earthquake (Mw9.0) had never been imagined.
1

Senior Researcher, Earthquake Disaster Prevention Division, National Institute forLand and Infrastructure Management (NILIM)䋭107䋭

In this paper, effects of long duration seismic motions on earthquake responseof highway bridges are investigated using the strong motion records from the 2003 off
Tokachi and the 2010 Maule earthquakes.
Strong Motion Records and Adjusted Waveforms
Table 1 lists 10 observation stations where strong motion records were obtainedby the Department of Geophysics, the University of Chile, during the 2010 Maule
earthquake. The digital data were downloaded from its website(http://ssn.dgf.uchile.cl). Locations of these observation stations are shown in Figure
1 with the epicenter and surface projection of the source fault. The strong motion
recorded at CCSP (Figure 2) has the largest PGA and the longest duration. In thispaper, the duration is defined as the time between first and last moments when
amplitudes exceed 50 [cm/s2]. The durations of NS, EW, and UD components of thestrong motion at CCSP are 151[s], 152[s], and 122 [s], respectively.
The acceleration response spectrum of EW component of the strong motion at
CCSP was adjusted to target response spectra by spectral fitting. The target response
spectra are set as shown in Figure 3. Type I, II, and III grounds are stiff, medium, andsoft soil conditions, respectively. Since there was no available information about thesoil condition at CCSP, three acceleration waveforms that represent Type I, II, and III
grounds were produced as shown in Figure 4; the acceleration response spectra of thesewaveforms were adjusted to the target response spectra (Figure 3) by spectral fitting.
The durations of the waveforms produced here are very close to the duration of theoriginal waveform (152 [s]).
Acceleration response spectra of strong motion recorded during the 2003 off
Tokachi earthquake were also adjusted to the same target response spectra by spectral
fitting. The strong motion records obtained at UKE (Urakawa-Efue), CKB(Chokubetsu), and TCS (Taikicho-Seika) stations were chosen to represent Type I, II,
and III grounds, respectively. The original and adjusted waveforms are shown in
Figures 5, 6, and 7. The durations of the adjusted waveforms are 75[s], 78[s], and 96[s],which are more than 50[s] shorter than those produced from the strong motion during
the 2010 Maule earthquake.
Analytical Models of Highway Bridges
Analytical model of highway bridges and nonlinear models for plastic hinge
section of RC piers and seismic isolation bearings are shown in Figures 8 and 9.
Rubber bearings, seismic isolation bearings, and fixed bearings were chosen for TypeI, II, and III grounds, respectively. Spread foundation was chosen for Type I ground,while pile foundation was chosen for Type II and II grounds. All three analyticalmodels were designed under the current seismic design specifications and theirfundamental natural periods are 1.25[s], 1.15[s], and 0.71[s] for Type I, II, and III
grounds, respectively.

䋭108䋭

Earthquake Response of Highway Bridges
Figure 10 shows hysteretic force-displacement response of the analytical model
subjected to the long duration seismic motions produced from the strong motion
records obtained during the 2003 off Tokachi and the 2010 Maule earthquakes. The
amplitudes of the adjusted waveforms were magnified to 1.2 times for seismic input tocheck nonlinear response of the model more clearly. We can see that peak response
displacements of the bridge models due to the adjusted waveforms from the 2003 off
Tokachi earthquake are as large as or larger than those from the 2010 Maule
earthquake.
Hysteretic force-displacement response of the seismic isolation bearings,
adopted for Type II ground, subjected to the long duration seismic motions from thetwo earthquakes are compared in Figure 11. The amplitudes of the adjusted waveforms
were magnified to 1.2 times for seismic input as well. The peak response
displacements of the bearings are not much different and do not exceed 250% shear
strain.
Table 2 summarizes residual displacements of the analytical models subjectedto the long duration seismic motions. The amplitudes of the adjusted waveforms weremagnified to 1.2 times and 1.4 times for seismic input. It was found that the residual
displacements vary case by case and that seismic motions with longer duration are notalways more severe in terms of the residual displacement.
Conclusions
In this paper, strong motion records obtained during the 2003 off Tokachi
earthquake (Mw8.0) and the 2010 Maule earthquake (Mw8.8) are used to investigate
effects of duration of seismic motion on earthquake response of highway bridges. We
could not find notable difference between the responses, i.e. the peak response
displacements of RC piers and seismic isolation bearings and the residual displacement,
excited by the seismic motions from the two earthquakes though there are differencesof more than 50 [s] between their durations. Further investigations will be carried out
using the abundant strong motion records obtained during the 2011 off the Pacific coastof Tohoku earthquake.
References
Japan Road Association (2002): Specifications for highway bridges, Part V: Seismic design.

䋭109䋭㪘㪺㪺㪼㫃㪼㫉㪸㫋㫀㫆㫅 㪲㪺㫄㪆㫊㪆㫊㪴

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㪫㫀㫄㪼 㪲㫊㪴㪈㪌㪇㪉㪇㪇(c) UD component (duration: 122[s])Figure 2 Strong motion recorded at CCSP during the 2010 Maule earthquake.

䋭111䋭㪘㪺㪺㪼㫃㪼㫉㪸㫋㫀㫆㫅 㪲㪺㫄㪆㫊㪆㫊㪴

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㪈㪌㪇㪉㪇㪇㪘㪺㪺㪼㫃㪼㫉㪸㫋㫀㫆㫅 㪲㪺㫄㪆㫊㪆㫊㪴(c) Type III ground (duration: 152[s])Figure 4 Acceleration waveforms produced by spectral fitting from EW component of
the strong motion at CCSP. The acceleration response spectra of these waveforms
were adjusted to the target spectra shown in Figure 3.
㪏㪇㪇
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㪈㪌㪇㪉㪇㪇㪘㪺㪺㪼㫃㪼㫉㪸㫋㫀㫆㫅 㪲㪺㫄㪆㫊㪆㫊㪴(a) Original waveform (duration: 46[s])
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㪫㫀㫄㪼 㪲㫊㪴㪈㪌㪇㪉㪇㪇(b) Adjusted waveform (duration: 75[s])Figure 5 Original (a) and adjusted (b) waveforms for Type I ground. The originalstrong motion is observed during the 2003 off Tokachi earthquake at UKE(Urakawa-Efue) station. Adjusted waveform was produced by spectral fitting

䋭113䋭㪘㪺㪺㪼㫃㪼㫉㪸㫋㫀㫆㫅 㪲㪺㫄㪆㫊㪆㫊㪴

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㪫㫀㫄㪼 㪲㫊㪴㪈㪌㪇㪉㪇㪇(b) Adjusted waveform (duration: 78[s])Figure 6 Original (a) and adjusted (b) waveforms for Type II ground. The originalstrong motion is observed during the 2003 off Tokachi earthquake at CKB(Chokubetsu) station. Adjusted waveform was produced by spectral fitting

䋭114䋭
Response of the seismic isolation bearings
Max. and min. response displacements

Horizontal force [kN]

Horizontal force [kN]

Yielding-250% shear strain

Max: 132mm
Min: -146mm

















Max: 155mm
Min: -132mm






Horizontal displacement [mm](a) CKB-Tokachi









Horizontal displacement [mm](b) CCSP-Maule

Figure 11 Hysteretic force-displacement response of the seismic isolation bearings,
adopted for Type II ground, subjected to the long duration seismic waves. The
amplitudes of the adjusted waveforms were magnified to 1.2 times for seismic input.

Table 2 Residual displacements of the analytical models subjected to the long duration
seismic motions. The amplitudes of the adjusted waveforms were magnified to 1.2
times and 1.4 times for seismic input.
GroundFundamental
natural period

Type I

1.25 [s]Type II

1.15 [s]Type III

0.71 [s]Strong motion record
UKE-Tokachi
CCSP-Maule
CKB-Tokachi
CCSP-MauleTCS-Tokachi
CCSP-Maule䋭118䋭

Residual displacement [mm]1.2 times
1.4 times
11
34
17
39
73
82
1
2
57
156
100
57

EXPERIMENTAL INVESTIGATION OF INFLUENCE OF LIVE LOAD ON
SEISMIC RESPONSE OF A HORIZONTALLY CURVED BRIDGE
Hartanto Wibowo1, Danielle M. Smith1, Ian G. Buckle2, and David H. Sanders3

Abstract
Little is understood about the dynamic interaction between heavy vehicles andbridge systems during strong earthquakes. An experimental study was therefore
undertaken to investigate this phenomenon. Six, full-scale, pickup trucks were placedon a 3-span, large-scale bridge model, which was supported on the 4 shake tables at the
NEES Equipment Site at the University of Nevada Reno. Comparison of behavior
when the same model was tested without trucks showed their presence to have abeneficial effect up to a level of shaking defined by the Design Earthquake, and an
adverse effect for shaking greater than the Design Earthquake. This bimodal result has
been reported by other researchers and confirms the difficulty of isolating andquantifying the critical parameters that govern response.

Introduction
Design procedures for earthquake-resistant bridges in most countries do not
require the simultaneous presence of live load and earthquake load to be considered.
This decision is based on two major assumptions. First, it is assumed the full design live
load will not be on the bridge at the time of the design earthquake, and second, theseismic response of a bridge is dominated by its dead load and live load inertial effects
are negligible by comparison. However for bridges in urban areas where congestion is afrequent occurrence, some fraction of the design live load (usually 50%) is now
recommended to be included with the dead load when computing gravity load effects(AASHTO, 2010). But this recommendation applies only to gravity load effects and not
to inertial effects.
The omission of inertial effects in design is the result of a prevailing attitude
that the suspension system of a heavy vehicle acts as a tuned mass damper and reduces
the motion in the bridge. It is therefore believed to be conservative to ignore theseeffects. But in fact little is understood about the dynamic interaction between heavy
vehicles and bridge systems during strong shaking and there is no hard evidence thatthe tuned mass damper model is universally applicable. It is equally possible that theadded weight increases the inertial loads in the bridge and the corresponding
displacements and forces.
Currently, very little research has been conducted to resolve this issue, and thefirst step is to understand and model vehicle suspension systems and their interaction
1

Graduate Research Assistant, Department of Civil and Environmental Engineering
Foundation Professor and Director of Center for Civil Engineering Earthquake Research
3
Professor, Department of Civil and Environmental Engineering, University of Nevada, Reno, MS 0258,
Reno, NV, 89557
2

䋭119䋭

with the bridge structural systems. Then these numerical models need to be calibrated
against experimental work, and finally the validated models should be used to study
bridge-vehicle interaction during earthquake ground shaking.
This paper describes an experimental study to calibrate a previously developed
numerical model and give some insight into the circumstances leading to beneficial andadverse behavior of live load during an earthquake. This study is part of a larger project
involving the development of a set of findings and recommendations concerning theeffect of live load on seismic response and how these effects may be included in theseismic analysis and design of bridges in the future.

Literature Review
Most of the research reported in the literature to date on the effect of live load isrelated to the calculation of impact factors (vehicle dynamic load allowance) for gravity
load design. It appears that very little work has been done on the influence of live loadon seismic response of bridges. Also, experimental tests on large-scale bridges to study
the effects of vehicle-bridge interaction under seismic load do not appear to have beenpreviously carried out.
A study by Sugiyama et al. (1990) used a single degree-of-freedom vehicle
system that can model rolling in the transverse direction and pitching in the longitudinal
direction, but the properties are not given. The bridge was idealized as a nine-masssystem with transverse and rotation inertia connected by linear springs and damper
elements. A vibration test is reported on an existing steel girder bridge with and without
trucks in the longitudinal and transverse directions to verify the results. In the test, twolarge trucks were parked facing the same direction on a portion of an existing off ramp
whose girders were vibrated with an electro-hydraulic exciter. The bridge was tested
with the vehicles empty and loaded to various capacities. The results show that the
dynamic effect of the vehicle is more dominant in the transverse direction and thevehicle tends to reduce the response of the bridge. The authors also note that as theexciting force level increases, the effects of nonlinearity become more apparent since
the dynamic characteristics of the vehicle itself are nonlinear. These results arecorroborated by Kameda et al. (1992) who used a 5 degree-of-freedom model in theirstudy. These authors state that the vehicle tends to increase the bridge response whenthe vehicle is in the in-phase mode with the bridge and decrease the bridge responsewhen it is in the out-of-phase mode. Moreover, they also concluded that the ratio of thefundamental frequency of the bridge to the vehicle plays an important role for the
response of the bridge.
Another study of the seismic response of a bridge with live load was done by
Kawatani et al. (2007). These authors analyzed the seismic response of a steel plate
girder bridge under vehicle loadings during earthquakes. The vehicles were modeledwith 12 degrees-of-freedom that took sway, yaw, bounce, pitch, and roll into account.
The observations from the numerical analysis showed that heavy vehicles, acting as adynamic system, can reduce the seismic response of bridges under a ground motion

䋭120䋭

with low frequency characteristics, but the vehicles have the opposite effect andslightly amplify the seismic response of the bridge under high frequency groundmotions.
Kawashima et al. (1994) and Otsuka et al. (1999) performed a series of studies
to determine the effect of live load on a bridge when combined with seismic load. The
study modeled a two-span simply supported girder bridge with a mix of ordinary cars,
modeled as additional dead load, and large trucks, each modeled with 5
degrees-of-freedom. The bridge was only analyzed in the transverse direction because
it was estimated that the deck response would be significantly affected by the rolling of
the large trucks. The studies found that the displacement response of the girders
increased by 10% when the live load was included; ductility demand at the bottom of
the column also increased by 10% with live load on the bridge. This study concluded
that these increases were not enough to be significant and that existing safety factors
should be adequate to cover these effects. It was also concluded that the increase in
response was due to the increase in weight. However, the effect of the large trucks was
not just to increase the dead weight, they also behaved as a mass damper.
Scott (2010) developed a simplified modeling approach for dynamic analyses toaccount for combined live load and seismic load. It is shown that for short-span bridges,
the displacement responses are mainly due to the fundamental bridge mode. In
addition, for long-span bridges, vehicle speed had small influence on the displacement
and acceleration responses of the bridge.
A recent study on the effects of live load a highway bridge under a moderate
earthquake in the horizontal and vertical directions is reported by Kim et al. (2011).
The study concluded that the seismic response of the bridge is amplified when thevehicle is considered as merely additional gravity load or mass and the amplification isdependent on the relationship between the fundamental frequency of the bridge and the
response spectrum of the ground motion. However, when the vehicle is considered as
dynamic or mass-spring-damper system, which is more realistic, the dynamic effect ofthe vehicle is greater than its gravity load effect and thus it reduces the seismic
response. In addition, the study also showed that the effect of a moving vehicle ascompared to a stationary vehicle is negligible, and it is sufficient to model the vehicle asstationary for these studies.

Bridge Model
A three-span, curved bridge model was tested in the Large-Scale Structures
Laboratory at University of Nevada, Reno. This 2/5-scale model has a steel plate girder
superstructure, single-column reinforced concrete substructures, and seat-type
abutments. Overall dimensions are shown in Table 1.
The bridge model has a total length of 145 ft, a total width of 12 ft, andsubtended angle of 104° as shown in Figures 1 and 2. Each bent has a single circularcolumn. The column height is 7 ft - 8 in with a diameter of 24 in. The superstructure is䋭121䋭

a three-span, three-girder steel bridge with concrete deck. The detail of thesuperstructure and the column can be seen in Figure 3. The superstructure is supported
by fixed (rotation-only) pot bearings at the bent locations and slider bearings at theabutments. Moreover, shear keys are provided at the abutments to restrain movement in
the radial direction during small amplitude earthquakes, but are designed to fail at
higher events to protect the abutment foundations against damage.
The prototype bridge was designed for a site in Seismic Zone 3 (AASHTO
2010) with a 1,000-year spectral acceleration at 1.0 sec (S1) of 0.4 g. Under this Design
Earthquake (DE), the bridge is expected to be damaged but not collapse. The record
selected as the input motion for the experimental studies was the Sylmar record fromthe 1994 Northridge Earthquake near Los Angeles, scaled to have the same spectral
acceleration at 1.0 sec. A scale factor of 0.475 was therefore applied to both the NS andEW time histories of ground acceleration from this station.
Table 1. Bridge Geometry Summary.
Parameter
Total Length
Span Lengths
Radius at Centerline
Subtended angle
Total Width
Girder Spacing
Total Superstructure Depth
Column Height
Column Diameter

Prototype362’-6”105’-0”, 152’-6”, 105’-0”
200’-0”104° (1.8 rad)30’-0”
11’-3”
6’-6.125”19’-2’
5’-0”

Model145’-0”42’-0”, 61’-0”, 42’-0”
80’-0”
104° (1.8 rad)12’-0”
4’-6”
2’-7.25”
7’-8”
2’-0”

Test Vehicles
The vehicles used in these experiments were six Ford F-250 trucks, eachweighing 10,000 lb (10 kip). Dynamic properties of a typical truck were found by shake
table testing using the 6 degree-of-freedom shake table in the Structures Laboratory.
The ratio of the total vehicle weight to the superstructure weight is around 22%. The
rationale of selecting the vehicle is discussed below.

䋭122䋭

(a)

(b)
Figure 3. Typical Superstructure and Column Details.
The starting point for selection of the test vehicle was the H-20 truck from the
Caltrans Bridge Design Specifications. This truck is a two-axle vehicle weighing 40
kips (8 kips on the front axle and 32 kips on the rear axle) with a 14 ft wheel base. For a
0.4-scale model, the model truck would have a wheel base of 5.6 ft, be 2.4 ft wide, andweigh 6.4 kip. Since such a vehicle would most likely have to be custom-built, thedecision was made to select from commercially available vehicles. The closest possible
vehicle to match the modeling requirements was found to be the Ford F-250. Although
the similitude requirements are not fully satisfied, the dynamic properties of the chosen
vehicle can produce similar effects to those of the target vehicle.

䋭124䋭

Table 2. Ford F-250 Dimensions and Weight Ratings
Parameter
Overall Length
Overall Width
Overall Height
Wheel Base Length
Ground Clearance
Curb Weight
Gross Vehicle Weight Rating
Max Allowable Payload

Value
247 in68 in
80 in
156 in7.9 in6.7 kip
10 kip2.3 kip

Experimental Setup
The bridge model was assembled on the four shake tables in the Large-Scale
Structures Laboratory and the vehicles positioned on the deck as shown in Figure 4.
Instrumentation has been installed on the columns, bridge girders, and trucks to gather
response data during testing. The types of instruments range from strain gauges on thecolumn rebar, string pots on the bridge girders and trucks (to measure displacements),
and accelerometers on the bridge deck and trucks (to measure accelerations). During
the experiment, 383 data acquisition channels were used.
The test protocol followed for this experiment started with 10% of the DE andthen the motion was increased in successive increments to 20%, 50%, 75%, 100%,150%, 200%, 250%, 300%, and 350% of the DE. Before each run, a series of white
noise excitations were run to characterize the system’s dynamic properties.

Figure 4. Bridge Model with Live Load (Courtesy of M. Wolterbeek, 2011)䋭125䋭Preliminary Results
One of the parameters that may be used to quantify the effect of live load is thecolumn displacement. Figures 5 and 6 show the north and south column displacementswith and without live load under 75% and 100% of DE, respectively. It is shown thatfor these two runs, the maximum displacement is less when live load is present. It isalso important to note that during the no-live load case, the shear keys at the abutment
failed during the 75% DE run, whereas it took a stronger ground motion (100% DE) to
fail these keys when live load was present, i.e. the live load reduced the forces in theshear keys at the same level of excitation. This shows that at these levels of shaking, theexistence of live load caused less demand in the column and reduced the radial shear
forces at the abutments. The damage in the column was also found to be minor and not
as severe as for the no-live load case.
On the other hand, observations from the higher amplitude runs, after the shear
keys at the abutments had failed, show a different result. Figures 7 and 8 show thedisplacements in the north and south columns with and without live load after 250%
and 300% of DE, respectively. It is seen at these levels of shaking (and after the keys
had failed), the live load exercises the columns to a greater extent and the maximum
displacements at the top of the columns became closer to the no-live load case. It is also
seen that the residual displacements in the columns for the live load case are about
double those without live load. These larger residual displacements indicate greater
distress to the columns, and especially the south column, due to the presence of the live
load.
Concluding Remarks
A recent experiment was conducted in the Large-Scale Structures Laboratory atUniversity of Nevada, Reno, to study the effects of live load on a 0.4-scale horizontally
curved bridge model. From the experimental results for the column displacements andradial shears forces at the abutments, with and without live load, some preliminary
conclusions can be drawn. In lower amplitude motions, when the shear keys were stillintact, live load gave a beneficial effect. In higher amplitude motions, after theabutments were free to move, live load gave an adverse effect. It is not known at thisstage whether this reversal in effect is due to (1) the deteriorating nature of the bridge
under increasing levels of shaking and thus a changing vehicle-to-bridge frequency
ratio, or (2) the changed configuration of the bridge when the abutments were releasedin the radial direction after the shear keys failed, or 3) both of the above. Studies arecontinuing to better understand this phenomenon.
Acknowledgments
This project is funded by California Department of Transportation undercontract number 59A0695 and Federal Highway Administration under contract number
DTFH61-07-C-00031. The experimental study was conducted at the George E. Brown,
Jr. Network for Earthquake Engineering Simulation (NEES) Equipment Site located at
University Nevada, Reno. The use of NEES shake tables at University of Nevada, Reno
was provided under a Shared-Use Agreement with NEEScomm.

䋭126䋭References
AASHTO. (2010). AASHTO LRFD Bridge Design Specifications, Customary U.S.
Units, 5th Edition. Washington, DC: AASHTO.
Kameda, H., Murono, Y., Maekawa, Y., and Sasaki, N. (1992). Dynamic
Structure-Vehicle Interaction for Seismic Load Evaluation of Highway Bridges.
Proceedings of the Tenth World Conference on Earthquake Engineering, Madrid,
Spain, July 19-24, pp. 4861-4866.
Kawashima, K., Unjoh, S., and Mukai, H. (1994). Research on Live Load Effect toEarthquake Response of Bridges (Part 1). PWRI Research Report No. 3316, 59 pp. (InJapanese)
Kawatani, M., Kim, C. W., and Yasui, K. (2007). Seismic Response of a Highway
Bridge under Traffic Loadings. Proceedings of Pacific Structural Steel Conference
2007: Steel Structures in Natural Hazards, Wairakei, New Zealand, March 13-16, pp.
183-188.
Kim, C. W., Kawatani, M., Konaka, S., and Kitaura, R. (2011). Seismic Responses of a
Highway Viaduct Considering Vehicles of Design Live Load as Dynamic System
during Moderate Earthquakes. Structure and infrastructure Engineering, 7 (7-8), pp.
523-534.
Otsuka, H., Unjoh, S., and Mukai, H. (1999). Research on Live Load Effect toEarthquake Response of Bridges (Part 2). PWRI Research Report No. 3355, 40 pp. (InJapanese)
Scott, M. H. (2010). Combined Seismic Plus Live Load Analysis of Highway Bridges,
Report OTREC-RR-09-261. Portland: Oregon Transportation Research and Education
Consortium.
Sugiyama, I., Kameda, H., Sasaki, N., and Kawakita, S. (1990). Dynamic
Structure-Vehicle Interaction of Highway Bridges and Its Implication to Seismic
Design. Proceedings of the 6th U.S.-Japan Bridge Engineering Workshop, Lake Tahoe,
NV, May 7-8, pp. 379-392.

Conversion Table
From
in
ft
lb

To
mm
mm
kg

䋭131䋭

Multiply by
25.4
304.8
0.45

䋭132䋭Discussion Method by using Dynamic Analysis for Quakeproofing Long-Span Bridges
Manager, Bridge Section, Road Department.
Deputy Manager, Bridge Section, Road Department
Assistant Manager, Bridge Section, Road Department
Bridge Section, Road Department

Takumi Nishikawa
Mutsuji Kawakami
Tatsumi Ujie
Shunji Nagahashi

1. Introduction
The capital in Western Japan, Osaka City, is the second largest city in Japan, with
only Tokyo surpassing it, and is the center for finance and culture. Osaka City is known as the“City of water”, and its bridges have been loved by the Japanese, down through the ages.
Currently, the Construction Section of Osaka City manages 764 bridges, which includes about
720,000 square meters of bridge. There are a great variety of bridges, including continuous
viaducts where 100,000 or more cars a day pass through, long-span bridges across large rivers(Picture 1), and bridges (Picture 2) that have been loved by the locals throughout history andare closely connected to the lives of the average person.

Figure 1 Location of Osaka City

Picture 1 Continuous viaduct(New Midosuji line)

Picture 2 Historical bridge(Namba Bridge)

䋭133䋭4. Survey on earthquake resistance of the current reinforced concrete bridge support
The basis of earthquake-resistant reinforcement for a reinforced concrete bridge
support is to change the failure mode to a flexure failure mode on the bridge support
foundation where toughness is expected, while avoiding shear failures which show drastic
failure on the bridge support.
The method described in “Existing bridge earthquake resistant reinforcement
method cases” 4) published by the Japan Bridge Engineering Center was used to evaluate
the damage on the cut-off area of longitudinal reinforcing steel. If the reinforcement of thebridge support foundation was implemented on a large scale, and it was difficult to change a
failure mode to a flexure failure mode of the bridge support foundation, we surveyed thatthe plasticity rate of the cut-off area was to the extent that the proof strength will not lower(the maximum response yield curvature of the cut off area should be twice the initial yield
curvature or less).
In order to confirm safety, the response rotation angles of the plastic hinges, theresponse shear force of the bridge supports, and the values of the residual displacement of
the bridge supports, acquired from dynamic analyses were surveyed. Concretely, we
surveyed that the maximum response rotation angle of the plastic hinges shall be theallowed rotation angle specified in Specification for Highway Bridges with Commentary,
PART V:SEISMIC DESIGN Section 10.2, or smaller, the maxim response shear force ofthe bridge supports shall be shear force of the reinforced concrete bridge support or lower,
and the residual displacements of the bridge supports calculated using the maximum
response plastic rate shall be the allowed residual displacement (1/100 of bridge support
height) or lower. As the bridge support is a wall type bridge support, the effects of the deep
beams are considerate for calculation of the shear proof strength at the transversal direction
to the bridge axis.
Earthquake resistance of a structure with the current bearing support condition(Table 2 CASE 1) being maintained was surveyed in the case of a ground motion at level 2.
The bridge support foundation and cut-off area of P2, P4, P5 and P7 bridge supports lacked
flexural and shear proof strength to the bridge axis direction (Table 3 CASE 1). P1 bridge
support showed an adequate earthquake resistance. Although the cut-off area of P3 and P6
bridge supports reached to the plastic field, that range was only to the extent in which proof
strength did not decrease.
All bridge supports showed the shear failure type to the transverse direction to thebridge axis. The shear proof strength calculated in consideration of the effects of the deep
beams exceeded the maximum response shear force, and met the required
earthquake-resistant strength. Therefore, if the bridge supports were mechanicallyreinforced against earthquake loads acting on each support, while the current bearing
support conditions were maintained, the reinforcement of the cut-off area of P2, P4, P5 andP7 bridge supports, and the shear reinforcement and flexural reinforcement of thefoundation would be necessary. Especially, the flexural reinforcement of the bridge support

䋭138䋭

foundation should be large scale reinforcements in which longitudinal reinforcing steel isestablished to the footing.
5. Discussion on earthquake-resistant reinforcement
5-1. Determination of the policy for reinforcement
For discussion of the whole structure of the Bridge, under the conditions ofbearing supports indicated in Table 2, a bridge support reinforcement method where
reinforcement against earthquake loads acting on the bridge support was implemented(CASE 1), a seismic isolation method which makes vibration periods of the structures
longer and enhances attenuation (CASE 2) and a seismic response control method which
enhances attenuation and distribution of earthquake loads (CASE 3) were weighed. Table 3
shows the results of the need of bridge support reinforcement.
In CASE 2, the natural period was lengthened by using seismic isolation bearing
supports, such as a laminated rubber bearing, and the inertial force was expected to be
reduced by improving the attenuation performance. Reinforcement of P1 and P6 bridge
supports was necessary, as well as reinforcement of P2, P4 and P5 bridge supports, and thescales of flexural reinforcement, shear reinforcement and the cut-off reinforcement werelarger, resulting in the construction costs being high. The traveling distances of the beam on
A1 abutment, P6 bridge support and A2 abutment were greater than those of the expansion
gap, resulting in a collision of the beams.
CASE 3 was a construction method which controls deformation of the entire
bridge and attenuates the inertia force transmitted to the bridge support by installing
dampers that deforms during an earthquake and absorbs the energy around the bearing
supports. The seismic response control dampers were located at A1 and P2 supporting
points of the 4 span continuous beam ranging from A1 to P4, P4 supporting point of theNielsen-Lohse beam ranging from P4 to P5, P5 supporting point of the simple beamranging from P5 to P6, P7 supporting point of the 2 span continuous beam ranging P6 to A2,
and A2 supporting point. In this structure, the inertia force transmitted to the substructure
dramatically decreased, shear and cut-off reinforcements were necessary only for P2, P4
and P5 bridge supports. As flexural reinforcement (establishment of longitudinal reinforcing
steel to the foundation) was unnecessary, the earthquake loads to the foundation did notincrease. A travelling distance of the beam at staggered areas was shorter than the expansion
gap with the adjacent beam, and beams did not collide with each other.
As a result of a comparison of the construction costs of these three methods, we
decided on the earthquake-resistant measure based on CASE 3, which controlled
deformation of the entire bridge and attenuated the energy by installing seismic responsecontrol dampers. For this basic structure, construction methods for the reinforcement of thebridge supports were compared, types of seismic response control dampers were compared,
and shear plate type dampers and friction hysteretic dampers were designed.

䋭139䋭

Table 2 Bearing support conditions

Table 3 List of need of bridge support reinforcements

5-2. Bridge support reinforcement
A PC confined method was adopted for reinforcement of the bridge support at
the low-water channel of a river. The PC confined method enables improvement of thehorizontal proof strength and toughness during an earthquake by prestressing the PC steel
wires inserted into the sheath in the PC precast panel after the panel is installed around thebridge supports, as well as improvement of work efficiency at the site, by using precast
panels manufactured at a plant.
A reinforced concrete lining method, a carbon fiber material lining method, and asteel plate lining method were compared and studied for the reinforcement of the bridge
support at the high-water channel of a river. As the carbon fiber material lining method has
excellent work efficiency but is vulnerable to fires and shocks, this method still has
operation and maintenance problems. The least expensive method of these methods, thereinforced concreted lining method was adopted.
5-3. Seismic response control dampers
Seismic response control dampers have shear panel and friction hysteretic
dampers. The shear panel damper (Picture 5) absorbs earthquake energy and reduces the
cross-sectional force which transmits to the substructure due to the shear deformation of thepanels which are composed of low yield point steel. This damper normally works as a
fixation device and when a level 1 earthquake occurs. The shear panel yields with groundmotion due to an earthquake at level 2 and this damper absorbs earthquake energy withhysteresis attenuation of shear plastic deformation. As this damper normally works as a
fixation device, it was installed at one supporting point of each beam, so that the beam could
be elastic during temperature changes (P4 supporting point of the Nielsen-Lohse beam andP5 supporting point of the simple steel plate 2 main beam I beam).

䋭140䋭

6 Conclusion
For earthquake-resistant reinforcement, structures of a general reinforcement
method which reinforces each bridge support against cross-sectional force acting on thebridge support while the current bearing support conditions are maintained were discussed
as well as the seismic isolation construction method, and the earthquake response control
method. Accordingly, the optimal method was selected where a high attenuation device,
such as a seismic response control damper, could absorb the earthquake energy, reduce theearthquake loads transmitted to the bridge supports, and control the deformation of theentire bridge. Then, the friction hysteretic damper and shear panel damper that use a low
yield point steel were designed, in consideration of the structure characteristics of the Bridge.
Currently, the reinforcement construction of the Nagara Bridge has been conducted for earlycompletion. We received guidance from Hirokazu Iemura (professor emeritus at Kyoto
University) for this paper. We will express our sincere gratitude to him.
Reference
1) Japan Road Association: Specification for Highway Bridges with Commentary, PART V:
SEISMIC DESIGN, Published by Maruzen, 2002-3
2) Takeda, T., Sozen, M.A. and Nielsen, N. M: Reinforced Concrete Response to Simulated
Earthquakes, Journal Division, ASCE, Vol. 96, No. ST12, PP. 2557-2573, December,
1970
3) Osaka City Planning Section: Osaka City Civil Engineering and Architectural Buildings
Earthquake Measure Technology Conference Report, 1997-3
4) Japan Bridge Engineering Center: Existing bridge earthquake resistant reinforcement
method cases, 2005-4
5) Maruyama. T. Seismic Retrofitting for Bridges in Osaka City, Osaka and Its Technology,
Osaka Municipal Government, No.32, PP7-18, 1998.1

䋭142䋭

27th US-Japan
Bridge Engineering Workshop

Session4

Management and Maintenance1
Analysis Examples of Periodic Inspection Results of National Highway Bridges in Japan-Application of Degree of Damage for Element LevelBy Fumihiko Nomura, Masanori Okubo, and Takashi Tamakoshi
The Importance of Partnerships in the Implementation of Bridge Preservation Practices
By Peter J. Weykamp
Maintenance of ASR-affected Structures in Hokuriku Expressway, JapanBy Yoshifumi Adachi, Mamoru Moriyama, and Masahiro Nomura

Page Left Blank

䋭144䋭

Analysis Examples of Periodic Inspection Resultsof National Highway Bridges in Japan-Application of degree of damage for element levelFumihiko Nomura1, Masanori Okubo2, Takashi Tamakoshi3
Abstract
In order to have enormous highway bridge stocks in Japan already, it isimportant to save safe and smooth traffic networks and to have them maintained both
economically and rationally. Thus, it is a matter of urgent business to grasp andevaluate their states appropriately, and to develop well-planned maintenance methods.
National Institute for Land and Infrastructure Management (NILIM) conducts
feature arrangements and analyses of damage by statistical processing of national
highway bridges periodic inspection data, and provides the insights into the further
rationalization and standardization of their periodic inspection.
In this paper, the features of generation and progress of damage on bridges arereported based on analysis examples of element periodic inspection results, and the
direction of the goal of bridge maintenance in Japan is explained.
Introduction
Periodic inspection of national highway bridges in Japan based on ‘Bridge
Inspection Manual’ (Public Works Research Institute Document No.2651, July 1988)
had been conducted every 10 years, and that based on ‘Bridge Periodic Inspection
Manual’ (National Highway and Risk Management Division, March 2004) (hereinafterreferred to as ‘Periodic Inspection Manual’) has been being conducted within two
years for initial inspection, and every five years after that since 2004.
Data as objective facts for damage and deterioration obtained from inspections ismade use of future projections and trend analyses by grasping continuously as well as
be indispensable for cause estimation and evaluation of current performances. In order
to be used for statistical processing and quantitative projection, it is important that thedata is the objective one based on a uniform standard possible to be used for relative
comparison over time.
On the other hand, for road administrators who do not always have the knowledge
enough to judge effects on bridge performance, primary diagnosis as evaluation offunctional states of bridges is essential to take action to make appropriate decisions fortaking measures to traffic regulations, and repairs and retrofits. In other words, except
the facts as individual kinds of damage and damage progression, it is possible for roadadministrators to make responses to conduct the further surveys right after obtaining
measures for effects of damage on functional states of bridges. So, in case primary
diagnosis has no troubles, it is so difficult for them to have suspicions about the results.
1

Researcher, Bridge and Structural Division, Road Department, NILIM
Senior Researcher, Bridge and Structural Division, Road Department, NILIM
3
Head, Bridge and Structural Division, Road Department, NILIM
2

䋭145䋭

DAMAGE PROGRESSION
A tendency of damage progression was analyzed by calculating transition
probability of degree of damage (from ‘a’ to ‘e’) for each member and region in orderto grasp how the degree of damage will progress within 5 years by using two periodic
inspection results (hereinafter called ‘the new inspection’ for the most recent
inspection, and ‘the old inspection’ below) conducted to the same bridge in the
different time.
A group of data consisting of the most recent periodic inspection results basedon the manual and the results based on the manual of 1988, and a group of two data ofbridges conducting two periodic inspection based on the manual are used for the
analyses.
For the verification of the old and new inspection results, the results which
satisfy the following extracted conditions were regarded as the effective data.
i) Inspection interval is within five years.
ii) The old and new inspection data for the same elements exist.
iii) No repairs were done between the old inspection and the new inspection.
iv) Degree of damage for the new inspection is the same or the worse as comparedto that for the old.
The total is 5,109 bridges, and the objective bridges were extracted for eachanalysis.
In this chapter, analysis examples of steel corrosion and crack of slab
remarkable for features of damage for each member and region in the previous chapter
were indicated. The analysis contents are a transition probability made up the degreeof damage for the new inspection to the old one, Markov chain based on that, an
expected value for Markov chains every five years weighting each degree of damage(‘a=1.00’, ‘b=0.75’, ‘c=0.50’, ‘d=0.25’, and ‘e=0.00’), and standard deviation in someof these graph.
(1) Corrosion of steel plate girder bridges
For corrosion of main girder of steel plate girder bridges in a A & B type of
painting, Figure 16 shows transition probability, Markov chain, and expected values
for whole members (from the left to the right), Figure 17 for the end of the girders,
Figure 18 for the middle, Figure 19 for the outside girders, and Figure 20 for the insidegirder.
For the transition probability of the left figure, for example, 80% of the degreeof damage ‘a’ (no corrosion) for the old inspection keeps the same state, while most of
the rest of 20% progress the degree of damage ‘b’, others progress ‘c’, ‘d’, and/or ‘e’.
Thus, the damage progression depends on the bridges, and so it has a tendency to vary
to some extent.
In this way, features shown in the damage progression states were also shownin the damage occurrence states.
Moreover, the deterioration prediction distribution varies, however, theaccuracy improvement of the deterioration curve by each region in Figure 17 & 18

䋭153䋭

THE IMPORTANCE OF PARTNERSHIPS IN THE IMPLEMENTATION OF
BRIDGE PRESERVATION PRACTICES
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䋭165䋭

䋭166䋭

Maintenance of AS R-affected Structures in Hokuriku Expressway, JapanM amoru M oriyama1, M asahiro Nomura2

Abstract
Increasing amounts of deicing salts have been used to ensure the safety of roadsurfaces in Hokuriku district, Japan. Sodium chloride, which accounts for the majority
of deicers, has been causing damage to road concrete structures due to alkali-silica
reaction. This report describes the current state of maintenance and repair techniquesof these road structures affected by deicer-related alkali-silica reaction.
Introduction
Increasing amounts of deicing salts have been used to ensure safety of roadsurfaces in Hokuriku district, in Japan. The deicers used in this area have been mainlyof sodium chloride (NaCl) type. In addition, the structures alongside the Japan sea
coastline are subjected to monsoons, which increase its exposure to airborne chlorides
from the sea. Therefore, these natural conditions set the road structures of the Hokuriku
district in a permanent saline environment. As the effect of this chloride supply, thecases that alkali-silica reaction (ASR) deteriorates road structures have been increasing.
On the other hand, river gravels and river sands have been used as concrete aggregates fora long time and these aggregates have been found in seriously ASR-affected structures.
ASR has occurred mainly in structures primarily built from 1970 to 1985, a time whenhigh alkali cement had been produced in Japan. Over the course of time, reactive riversands and gravels may not be properly assessed since these aggregates contain a wide
variety of rock types, as well as different types and contents of reactive minerals.
Central Nippon Expressway Company has covered the sections from Kinomoto
IC (Shiga Prefecture) to Asahi IC (Toyama Prefecture) in Hokuriku Expressway andOyabe-tonami JCT (Toyama Prefecture) to Shirakawago IC (Gifu Prefecture) in
Tokai-Hokuriku Expressway, a total of 300km (9-39 years in service). There are 545
bridges, 1200 C-Boxes and 45 tunnels in these sections. ASR occurs in some of thesestructures and the application of efficient maintenance techniques have been put inplace.
This report presents the current state of inspection and survey of road structures
under saline environment in the Hokuriku district, the methods of diagnosis, repair andrepair prioritization of ASR damaged structures.
1

Central Nippon Expressway Company Limited Kanazawa Branch

2

Researcher, Central Nippon Highway Engineering Nagoya Company Limited Kanazawa Branch

䋭167䋭

The AS R Characteristics in Hokuriku district
The Hokuriku district is located approximately in the center part of the JapanSea coastline and is characterized by severe weather conditions: hot and humid summers
and cold and snowy winters (see Fig. 1㨪Fig. 3).
Figure 4 shows the map of deteriorated structures in the Hokuriku district. In theexpressways ASR occurs in almost all areas of Toyama Prefecture and parts of Ishikawa
and Fukui Prefectures. The reactive aggregates were confirmed as andesite, rhyolite andtuff, from volcanic rock deposits (see Fig. 5). Deterioration of road structures by ASR
has been often assessed as a combined effect of rainwater, road drainage and deicer
application (see Fig. 6).
Maintenance
The inspection of road structures is classified as regular inspection or detailed
inspection. Regular inspections are conducted once a year, while detailed inspections arecarried out in the span of 2 to 5 years. These inspections have been mainly visual. In
addition, continuous survey of the expansion behavior of the structures (by contact
JDXJH RU E\ ʌ JDXJH is conducted every year and followed by a thorough investigation.
Fig. 7 shows the flow of assessment of ASR-deteriorated structures, which has
been applied in reference to the case-studies shown in Fig. 6. In addition, the structure’s
database is comprised with the construction company, concrete plant, cement maker,
aggregate quarry or supplier once construction is elaborated. In the deterioration cases
where the assessment is inconclusive, ASR occurance is determined by a comparativeevaluation of the database of other structures supplied from the same concrete plant. If
it cannot provide a proper judgment, then a detailed investigation has to be carried out.
Detailed ASR investigation includes the mineralogical composition of coarse
aggregates, thin section observation with a polarizing microscope, alkali and chloride
contents in concrete, and the residual expansion of concrete cores drilled from thestructures. In particular, considering the feature of Hokuriku district, the residual
expansion test of concrete cores, which is a technique that conforms with the alkali
external supply, has been adopted . The test method is described in Table 1 and shown
through Fig.7㨪Fig. 11.
Repair and monitoring
Figure 12 shows the flow of repair method selection of ASR-deteriorated
structures. Since there are a considerable number of ASR deteriorated structures in the
Hokuriku district and, therefore, a need for reducingmaintenance costs, a ranking system
based on the visual inspection has been created for chosing the ASR repair method. For

䋭168䋭

ranking“A” structures, crack injection and surface coating is applied, including the timereference results of residual expansivity tests of concrete cores. For ranking “B or C”
structures, the road surface drainage countermeasures are applied, followed by close
observation of the improvements. Repair rankings are priority for importantintersections (such as national road and railway) and the areas affected by concrete
spalling or peeling off, works include section repair and coating with fiber sheets. In
addition, due to the severe environment conditions in the Hokuriku region, therehabilitation period has been set to about one or two year longto ensure quality control.
After the repairing is concluded, monitoring is applied for inspection purposes.
Focus point is the effect of road surface drainage. Also, a special attention is paid tobridge abutment front, where the water absorption from the back part often leads toswelling of the surface coating. The effective countermeasures for stopping water
migration from the back part of bridge abutments are a matter of serious concern.
Concluding Remarks
The extent of ASR deterioration of concrete structures in the Hokuriku district issignificant if compared to other areas of Japan. In some cases, fracture of reinforcingsteel
bars due to ASR expansion has been reported. The lack of appropriate repairing
countermeasures to stop ASR progress is still a matter of fact. If the repair occurs ASR,
there will be no drastic measures. At present, ASR in bridge superstructure has notreached deterioration levels which exert influence on the load bearing capacity. In the
future we will plan to continue the conduction of detailed inspections and surveys onthese ASR-affected structures.

Lithologicalcompositionof gravel
Alkali-silica
reactivity:

Alkali content
of concrete:

Table 1 Outline of ASR test methods
In regard to particles 5mm or more in diameter visible on thesurfaces of concrete cores, the lithological composition is
calculated by the point count method using a 5mm mesh.
Alkali-silica reactivity is judged by the presence of rims
and/or alkali-silica gel around aggregate particles. Thin
sections are prepared from some specimens for polarization
microscopy observation.
A 10-g sample ground to particles less than 150Pm in
diameter is added to 100ml of distilled water at 40qC andagitated for 30 min. The mixture is then filtered, and thealkali concentration of the filtrate is measured by atomic
absorption photometry, to calculate the Na2O equivalent(Na2O + 0.658 K2O).

䋭169䋭

Crack injection

Surface coating

Water jet

Section repair with shotcrete Execution of fiber sheet coating Completion of repairing
works
Fig.13 Maintenance of ASR-affected bridges

䋭174䋭

27th US-Japan
Bridge Engineering Workshop

Session5

Load and Strength Evaluation
Load-Carrying Capacity of Reinforced Concrete Beams with Adhesively Bonded Steel Plates
By Yoshiki Tanaka, Jun Murakoshi and Eiji Yoshida
Summary of NCHRP Research on Recalibration of the LRFR Load Factors in the AASHTO
Manual for Bridge Evaluation
By Waseem Dekelbab
Compressive Loading Test of Corroded Gusset Plate Connection in Steel Truss Bridge
By Jun Murakoshi, Naoki Toyama, Mamoru Sawada, Kentaro Arimura,
Lu Guo, Kuniei Nogami, Teruhiko Yoda, and Hideyuki Kasano

Page Left Blank

䋭176䋭

LOAD-CARRYING CAPACITY OF REINFORCED CONCRETE BEAMS WITHADHESIVELY BONDED STEEL PLATES
Yoshiki Tanaka1
Jun Murakoshi1
Eiji Yoshida1

Abstract
To evaluate the load-carrying capacity of reinforced concrete beams withexternally bonded steel plate, the influence of the bonded steel plates to the soffit on theshear strength of the beams, and the effect of shear strengthening using wing-type sidesteel plates for stocky concrete beams (broad and low web) have been examined. From theresults, it was found that the shear strength is properly evaluated by shear capacity
equations for non-plated reinforced concrete beams unless a crack develops at the edge ofthe steel plate, and that the side plate bonding is likely to have a potential to be effective instrengthening for the stocky beams without stirrup. Additionally, loading tests using two
75-year-old deteriorated reinforced concrete beams with adhesively bonded steel plates tothe soffit were carried out. The bonded steel plates no longer contributed to the load
carrying capacity after the joints between the steel plates failed due to debonding.

Introduction
In Japan, a number of the concrete girders and decks of existing highway bridges
were strengthened with adhesively bonded steel plates or fiber reinforced polymer (FRP)
sheets when design loads were changed or when the members deteriorated (Figs. 1 and 2).
For the efficient management of existing bridges, the load-carrying capacity of not only
conventional concrete beams but also concrete beams with the bonded steel plates should
be properly evaluated.
The late 1960s, several researchers dealt with the strengthening method that steel
plates were adhesively bonded with a two-part epoxy resin adhesive on the soffit ofreinforced concrete beams. In those days, because the fatigue deterioration of concrete
decks was frequently found on the Japanese highway bridges, experimental studies on thesteel plate bonding technique applying to the bridge decks were carried out.1-4 Then the
FRP sheets were not major materials for construction. The steel plate bonding immediately
became a major tool to improve the durability of bridge decks, having the advantage ofapplicability in service, and the efficient strengthening with the minimum change in
1

Public Works Research Institute, Japan

䋭177䋭

appearance. The strengthening method using FRP sheets for reinforced concrete wasdeveloped in early 1990s. Currently, the materials became used commonly to improve thedurability of bridge decks after a revision of the legal maximum weight for trucks from 20
tons to 25 tons in 1993, and the ductility of existing bridge piers after the great Hanshin
earthquake in 1995. Extensive research works were carried out not only in Japan, but also
in the world. Based on the results, several design guidelines for strengthening reinforced
concrete members using FRP have been established.5-8 On the other hand, the effects of thesteel plate bonding technique has not been so sufficiently identified as a guideline could beestablished. In a research program from FY2008 to 2010, the effect of shear strengthening
using the steel plate bonding on reinforced concrete beams was examined. This summary
report provides the outlines of three series of experimental studies focusing on; Series I:
the influence of the steel plate bonded to the soffit on the shear capacity of reinforced
concrete beams (further for the evaluation of plated concrete decks), Series II: the effect ofside steel plate bonding on the shear capacity of stocky reinforced concrete beams,9 andSeries III: the load-carrying capacity of two 75-year-old deteriorated concrete girders withthe steel plate bonded to the soffit.10
Series I: Influence of Steel Plate Bonded to the Soffit on Shear Capacity ofReinforced Concrete Beams
Test Program: A test scheme and details of specimens are shown in Fig. 3 and Table 1.
The height of the specimens, except Specimens A-7 and A-8, was set as 190 mm, referringto the thickness of the typical bridge decks built in 1960s. No stirrup was arranged exceptthat on the support points for arranging main reinforcing bars. Specimens A-1 to A-8 had
two longitudinal reinforcing bars. For Specimens A-4 to A-8, the steel plate with athickness of 4.5 mm was adhesively bonded on the soffit of concrete beams. The steel plate
and an adhesive layer for every beam were as wide as the beam with a width of 150 mm.
The thickness of the adhesive layer was 5 mm. The distance from the center of each thesupport point to the end of the steel plate xa was 80 mm, and that from the edge of a steel
plate for the support was 30 mm.
The test parameter of Specimens A-4 to A-6 is the ratio of shear span a to effective
depth ds, which excludes the dimension of the steel plate. The effective depth and shear
span of Specimens A-7 and A-8 ware larger than those of Specimen A-5 with the same a/ds
ratio. Control specimens A-1 to A-3 with no steel plate had different reinforcement ratios,
respectively.
Specimens B-4 to B-6, and B-8 had a main reinforcing bar and a doubly reinforcing
bar, the steel plate being bonded to the soffit. The thickness of the steel plate of Specimens
B-4, B-5, and B-6 was 4.5, 6, and 12 mm, respectively. Specimens B-4 to B-6 were simply
supported on the bonded steel plate, as the shear span was 400 mm. Then the bonded steel
plate had no curtailment near the supports. For Specimen B-8, the steel plate was bonded
after the concrete beam had been precracked under cyclic loading with a shear span of 500

䋭178䋭

mm. Specimen B-8 was monotonically loaded with a shear span of 1000 mm, having thecurtailment of the steel plate with a distance to the center of the support xa of 80 mm.
Plate bonding was carried out at three weeks or more after casting. Anchor holes
were drilled and mechanical steel anchors were installed on the soffit of concrete beams,
which was roughened with a hand grinder. The steel plate was supported with the anchors
as the gap to the surface of concrete was held with 5 mm spacers. Side openings ware
capped with epoxy resin putty. The epoxy resin adhesive was filled into the gap by
grouting. In Specimens A-4 to A-8, inorganic zinc-rich primer for protecting the steel plate
up to bonding remained similarly to practice. The steel plates for Specimens B-4 to B-6,
and B-8 were not coated.
Loading tests were conducted at a week or later after the grouting. The putty on thesides was ground; that at the ends of the steel plates remained. The tensile-shear bondstrengths of the epoxy resin adhesives were 12.3 MPa to 18.4 MPa.
Results: All test results are also shown in Table 1. Specimens A-4 to A-6 failed due tocracking at the end of steel plate (Fig. 4).The shear force at cracking at the end of the steel
plate Vp2 may be calculated based on a model provided by Tumialan et al.11 In the model,
the maximum (tensile) principal stress acting on the concrete near the end of the bonded
steel plate is estimated based on the approximate solution by Roberts12, being comparedwith the tensile strength of concrete. Although Tumialan et al.11 applied the modulus ofrupture fr to the threshold tensile strength, the splitting tensile strength ft is applied in thisreport, because it is more understandable for the model. Figure 5 shows the relationshipbetween the calculated shear force at the cracking Vp2 and the experimental results Vex
including the results of Specimens A-4 to A-6 and the other experimental results obtained
from previous research13-24 except the results of specimens with a ratio xa /a of more than
0.25. The calculated values Vp2 are somewhat larger than the experimental values when thevalue is larger. However, considering that the previous data contain the several unknown
parameters, it appears that the shear force at the cracking can be properly estimated.
Incidentally, the database based on the previous research shows that the results ofspecimens with a ratio xa /a of more than 0.25 showing debonding failure do not depend on
the value of Vp2, rather they mainly depend on the ultimate flexural strength of non-plated
reinforced concrete beams.
In Fig. 6a, the results of specimens exhibiting flexure-shear failure in concrete areshown in relation to the calculated shear strength Vsh based on the conventional empiricalequation25 for estimating a shear force at flexure-shear failure. The influence of the bonded
steel plates is estimated by substituting the dimensions containing the steel plate for the
calculation of the reinforcement ratio and the effective depth in the equation. The othertriangle symbols indicate the results of control specimens containing Specimens A-1 to
A-3 and the previous research related. For the results of specimens with a ratio a/d of 1.7
to 1.75, the other empirical equation for deep beams26 is applied. From the results, it was

䋭179䋭found that the shear capacity of the plated beams can be estimated by the conventional
equations. For reference, the calculated shear strength based on an equation in ACI318 (Eq.
11-5)27 with a factor of 1.3 is also compared with the test results except the deep beam in
Fig. 6b.
Series II: Effect of Side Steel Plate Bonding on Shear Capacity of Stocky Reinforced
Concrete Beams9
Test Program: Configurations of specimens and a test setup are shown in Figs. 7 to 8. Of
the same dimension of two reinforced concrete beams, Specimen S, of which wing-type
steel plates was adhesively bonded on the sides of web at both shear spans. The other beamwith no steel plate was named Specimen N. The concrete beams were designed as a simply
supported T-shaped beam with a stocky cross section consisting of a web width of 600 mm
and a height of beam of 850 mm, containing ten No.11 longitudinal reinforcing bars witha yield point of 534 MPa. For both specimens, four 9 mm dia. stirrups with a spacing of 250
mm were arranged at a shear span (right side in Fig. 8), no stirrup being at the other shear
span. Similarly to the old reinforced concrete bridges, all the stirrups were round steel bars
with a yield point of 292 MPa. The compressive cylinder strength of concrete was 33 MPa.
Four steel plates with a size of 2400 mm x 520 mm x 4.5 mm and a yield point of
360 MPa, coated by inorganic zinc-rich primer, were prepared. The procedure for bonding
was the same as Series I. The tensile-shear bond strength of the epoxy resin adhesive was
16.1 MPa. Twelve steel anchors with a size of M10 and an embedded length of 70 mm forsupporting each the steel plate at grouting were preinstalled before casting. The thicknessof adhesive of 5 mm was kept.
Two point loading were monotonically applied to the beams with a shear span of
2 m. Cracking inside the web concrete covered by the steel plates were monitored using
molded gauges and gauges on the stirrups. After shear failure occurred at the shear spanwith no stirrup, once unloaded. Then a vertical restrainer consisting of four tendon bars
with 13 mm dia. were installed at 1/3 of the shear span with no stirrup from the nearloading point. Reloading beyond the yielding of stirrups became available by using the
restrainer, although finally the flange yielded at the shear span with no stirrup.
Results: Relationships between load and deflection at midspan for both specimens areshown in Fig. 9. Cracks observed on the web at the first peak due to shear failure at theshear span with no stirrup are shown in Fig. 10. The load at each event in the loading tests
is summarized in Fig. 11. The load at diagonal cracking (detected with the molded gaugesin Specimen S) was attributed to neither the stirrup nor the bonded steel plate. For the web
without stirrup, the bonded steel plates showed an appreciable effect of improving theshear capacity, which was approximately 1.5 times as high as that of the non-plated
specimen.

䋭180䋭

The side steel plates of both shear spans were debonded at similar loads regardlessof stirrups, while the debonding of the steel plates on the web with stirrups did not
immediately expand compared with that on the web without stirrup. At the shear span withstirrups, even after debonding, the steel plates were to some extent likely to contribute to
the delay in failure. The stirrups, however, showed ductile behavior compared with anestimated level based on the modified truss theory despite the bonded steel plates.
Series III: Loading Tests of Old Concrete Beams with Steel Plate Bonding10
Outline of test beams: To examine the load-carrying capacity of old reinforced concretebeams with externally bonded steel plates to the soffit, six beams with a span length of 10
m were taken from the first and seventh spans of a concrete highway bridge built in 1935
(Fig. 1). In about 1981, steel plates were bonded to the soffit of most decks and beams inthe bridge. Because the significant deterioration of the decks and beams was found when
the administrator was considering the delivery of the bridge to the other small
administrator, eight deteriorated spans of the bridge were replaced in 2010. According to
the results of the investigation and observation before the removal, the degree of thedeterioration of eight spans differed widely from each other. On the beams at the first span,corrosion of steel plates, water leakage and cracks in concrete were widely observed. The
beams at the seventh span less deteriorated than that at the first span. Of six beams obtained,a beam from the seventh span and another beam from the first span, named Specimen S1and S2, respectively, were tested.
Observations and survey of materials: Cracks observed on the test beams before theloading test and each cross section cut after the test are shown in Fig. 12. Coating partiallycovering web concrete was removed for the observations. The cracks were widely
observed on the web of Beam S2. The crack depth, however, was found to be shallower
than that concerned. Innumerable horizontal cracks were found in the flanges of bothbeams. Hammer soundings detected the debonding area of the steel plates as shown in Fig.
12. From a partial dissection survey conducted after the loading tests, two 8 mm dia.
stirrups with a spacing of approximately 200 mm were found. The mechanical propertiesof concrete cores at uncracked parts, reinforcing bars and the steel plate taken from thetested beams are shown in Tables 2 to 3. It should be noted that the modulus of elasticityof the concrete cores taken from Beam S2 was considerably lower than that of Beam S1.
Test program: Two point loading were carried on the beams with a span length of 10
meters as the original span. Two strain gauges were mounted on the main reinforcing bars
at a loading point section, after the web concrete was dug as small as possible. Asphalt
surfacing remained at loading, because the removal would affect the cracked flange, and an
additional preparation for the surface would be required. The tests were carried out inwinter. For Beam 2, vertical restrainers were installed at the joints between the bonded
steel plates in order to temporarily attempt to mitigate the debonding of the splice plate,although no effect was found from the result.

䋭181䋭

Results: Relationships between load and deflection at midspan of both the old beams areshown in Fig. 13. In Table 4, the maximum loads are compared with calculated ultimate
loads, which were obtained using the measured dimensions and the properties of materials,
except the asphalt surfacing. The sign of debonding of splice plates in Beam S1 wasdetected at 400 kN by using LVDTs (Fig. 14a), and in Beam S2 at 320 kN. After the signof debonding in Beam S1, the load further increased up to the ultimate flexural capacity of
the conventional reinforced concrete beam with no deterioration, while the rigidity
significantly decreased. The flexural capacity of the deteriorated beam S2 was 8% lower
than that of Beam S1, being 5% lower than the estimated value. After the peak, yielding of
the main reinforcing bars was recognized by strain readings, and the flange was moderately
crushed (Fig. 14b).
Conclusions
1 - From the test results and the previous research, it was found that the shear capacity ofreinforced concrete beams with a steel plate bonded to the soffit is properly evaluated bysubstituting the dimension containing the steel plate for the calculation of thereinforcement ratio and the effective depth in the shear capacity equations for non-plated
reinforced concrete beams, unless a crack develops at the edge of the steel plate.
2 - The side plate bonding is likely to have a potential to be effective in strengthening forthe stocky reinforce concrete beams without stirrup. The effect of the side steel plates,
however, may be limited depending on the web reinforcement ratio in existing beams.
3 - In the results of loading tests using old reinforced concrete beams, the bonded steel
plates no longer contributed to the load carrying capacity after the steel splice plates failed
due to debonding.
Acknowledgments
The authors thank Nagano prefecture and engineers who assisted with the surveyof the old bridge.
References
1. Ohta, M., Aizawa, I., Momota, K., and Nakamura, T., Experimental Study on Strengthening Using Steel
Plate Bonding for Reinforced Concrete Decks, Tech. Note of PWRI, No. 681, April 1971. (in Japanese)
2. Ishitani, T., Higai, T., and Hosoda, K., Experimental Study on Strengthening Methods for Reinforced
Concrete Decks, -Steel Plate Bonding Technique-, Report of Japan Highway Corporation Laboratory,
pp.150-157, Dec. 1975. (in Japanese)
3. Iioka, Y., and Higai, T., Experimental Study on Strengthening Methods for Reinforced Concrete Decks,-Steel Plate Bonding Technique (Part 2)-, Report of Japan Highway Corporation Laboratory, pp.142-154,
Nov. 1976. (in Japanese)
4. Iioka, Y., Higai, T., and Hosoda, K., Experimental Study on Strengthening Methods for Reinforced
Concrete Decks, -Steel Plate Bonding Technique (Part 3)-, Report of Japan Highway Corporation
Laboratory, pp.143-151, Dec. 1977. (in Japanese)

䋭182䋭5. Recommendations for Upgrading of Concrete Structures with Use of Continuous Fiber Sheets, Concrete
Library 101, Japan Society of Civil Engineers, July, 2000. (in Japanese)
6. Mufti, A.A., et al., New Canadian Highway Bridge Design Code design provisions for fibre-reinforced
structures, Can, J, Civ. Eng., 34, pp.267-283, 2007.
7. ACI 440 Committee, Guide for the design and construction of externally bonded FRP systems forstrengthening concrete structures, ACI, July 2008.
8. Zureick, A., Ellingwood, B. R., Nowak, A. S., Mertz, D. R., and Triantafillou, T. C., Recommended Guide
Specification for the Design of externally Bonded FRP Systems for Repair and Strengthening of Concrete
Bridge Elements, NCHRP Report 655, TRB, 2010.
9. Yoshida, E., Murakoshi, J. and Tanaka, Y, Effect of Adhesively Bonded Steel Plates on Web
Reinforcement for Reinforced Concrete Beams, Proceedings of JCI, pp.703-708, 2010. (in Japanese)
10. Yoshida, E., Murakoshi, J. and Tanaka, Y, Structural Testing of Deteriorated Reinforced Concrete
Girders Strengthened by Externally Bonded Steel Plates, Proc. JSCE Annual meeting, V, pp.377-378,
2010. (in Japanese)
11. Tumialan, J. G., Belarbi, A., and Nanni, A., Reinforced Concrete Beams Strengthened with CFRP
Composites: Failure due to Concrete Cover Delamination, Center for Infrastructure Engineering Studies,
99-01, Department of Civil Engineering, Univ. of Missouri - Rolla, March 1999.
12. Roberts, T. M., Approximate analysis of shear and normal stress concentrations in the adhesive layer ofplated RC beams, Structural Engineer, 67(12), pp.229-233, 1989.
13. Jones, R., Swamy, R. N., Bloxham, J., and Bouderbalah, A., Composite behaviour of concrete beams withepoxy bonded external reinforcement, Int J Cem Compos, 2(2), pp.91-107, 1980.
14. Macdonald, M. D., The flexural performance of 3.5m concrete beams with various bonded external
reinforcements, TRRL Supplementary Report 728, 1982.
15. Jones, R., Swamy, R. N., and Ang, T. H., Under- and over-reinforced concrete beams with glued steel
plates, Int J Cem Compos Lightweight Concr, 4(1), pp.19-32, 1982.
16. Swamy, R. N., Jones, R., and Bloxham, J. W., Structural behaviour of reinforced concrete beams
strengthened by epoxy-bonded steel plates, Structural Engineer, 65A(2), pp.59-68, Feb. 1987.
17. Jones, R., Swamy, R. N., and Charif, A., Plate separation and anchorage of reinforced concrete beams
strengthened by epoxy-bonded steel plates, Structural Engineer, 66(5), pp.85-94, 1988. Correspondence
from Roberts, 67(12), pp.187-188, 1988.
18. Oehlers, D. J., and Moran, J. P., Premature failure of externally plated reinforced concrete beams, J. of
Structural Engineering, ASCE, 116(4), pp.978-995, April 1990.
19. Basunbul, I. A., Gubati, A. A., Al-Sulaimani, G. J., and Baluch, M. H., Repaired reinforced concretebeams, ACI Materials J., 87(4), pp.348-354, July-Aug. 1990.
20. Oehlers, D. J., Reinforced concrete beams with plates glued to their soffits, J. Structural Engineering,
ASCE, 118(8), pp.2023-2038, 1992.
21. Hussain, M. et al., Flexural behavior of precracked reinforced concrete beams strengthened externally bysteel plates, ACI Structural J., 92(1), pp.14-22, Jan.-Feb. 1995.
22. Baluch, M. H., Ziraba, Y. N., Azad, A. K., Sharif, A. M., Al-Sulaimani, G. J. and Basunbul, I. A., Shearstrength of plated RC beams, Mag. of Concrete Res., 47, No. 173, pp.369-374, 1995.
23. Swamy, R. N., Jones, R., and Charif, A., Contribution of externally bonded steel plate reinforcement tothe shear resistance of reinforced concrete beams, Repair and Strengthening of Concrete Members with
Adhesive Bonded Plates, SP-165, ACI, pp.1-24, 1996. (The session was held in 1992)
24. Sano, M., and Miura, T., A study on a design method for strengthening concrete members bysteel-plate-bonding, Proc. of JSCE, 550, pp.117-129, Nov. 1996. (in Japanese)
25. Okamura, H., and Higai, T., Proposed design equation for shear strength of reinforced concrete beams
without web reinforcement, Proc. of JSCE, 300, pp.131-141, Aug. 1980.
26. Niwa, J., Equation for shear strength of reinforced concrete deep beams based on FEM analysis, Proc. of
JCI 2nd Colloquium on Shear Analysis of RC Structures, pp.119-128, Oct. 1983. (in Japanese)
27. ACI 318 Building Code Requirements for Structural Concrete and Commentary, ACI, 2005.

䋭183䋭

Table 1 Details and test results of specimens in Series I
Specimen
A-1
A-2
A-3
A-4
A-5
A-6
A-7
A-8
B-4
B-5
B-6
B-8
Steel plate strengthening
No
No
No
Yes
Yes
Yes
Yes
Yes
Yes
Yes
Yes
Yes
Continuity of steel plate at support
No
No
No
No
No
Yes
Yes
Yes
No
㸫
㸫
㸫
Height of beam h, mm
190
190
190
190
190
190
250
330
190
190
190
190
Shear span a, mm
480
480
480
320
480
640
660
900
400
400
400 1000
Effective depth d, mm
160
160
160
183
183
183
237
315
189
191
197
189
(after strengthening)
a/d
3.0
3.0
3.0
1.7
2.6
3.5
2.8
2.9
2.1
2.1
2.0
5.3
Longitudinal reinforcing bars
2-#7 2-#5 2-#6 2-#5 2-#5 2-#5 2-#7 2-#8 1-#5 1-#5 1-#5 1-#5
Ratio of reinforcement before
3.23 1.66 2.39 1.66 1.66 1.66 2.35 2.25 0.83 0.83 0.83 0.83
strengthening, %Ratio of reinforcement after
3.90 3.90 3.90 4.07 3.57 3.09 3.83 6.77 3.09
strengthening, %Doubly reinforcing bars
1-#5 1-#5 1-#5 1-#5
Thickness of steel plate tsp, mm
4.5
4.5
4.5
4.5
4.5
4.5
6.0
12.0
4.5
Comp. cylinder strength of
27.7 28.0 28.5 29.4 28.6 29.7 29.8 30.6 27.8 27.8 27.8 28.3
concrete, MPa
Splitting tensile strength of
2.3
2.3
2.3
2.3
2.3
2.3
2.3
2.3
2.0
2.0
2.0
2.4
concrete, MPa
Yield point of reinforcing bars,
369
390
374
390
390
390
369
362
348
348
348
348
MPa
Yield point of steel plate, MPa
300
300
300
300
300
306
318
330
306
Diagonal cracking load Pcr,ex, kN
88
77
79
78
75
73
121
158
90
112
131
86
Maximum load Pmax,ex, kN
122
94
79
121
78
74
121
158
86
Type of failure
S
S
S
D
D
D
S
S
**
**
**
S†
Note: 1) Width of beams and plates : 150 mm, Thickness of adhesive layer : 5 mm2) **: Loading was stopped at diagonal crack load.
3) A steel plate was installed on the soffit of Specimen B-8, after initial cracks developed due to cyclic loading when theshear span was set as 500 mm.
4) S : shear failure, D : failure due to cracking at the edge of steel plate,† : Shear failure happened after the yielding of steel plate.

Fig. 4 Failure due to cracking at the edge ofthe bonded steel plate in Specimen A-6

䋭185䋭Summary of NCHRP Research on Recalibration of the LRFR Load Factors
in the AASHTO Manual for Bridge Evaluation
Waseem Dekelbab1
Abstract
This paper summarizes the findings of research conducted under NCHRP Project
20-07, Task 285 “Recalibration of the LRFR Load Factors in the AASHTO Manual for
Bridge Evaluation”. The objectives of the project were to (1) develop and recommend
reliability indices better aligned with current permit operations for routine and special
permit calibrations and (2) recalibrate LRFR live load factors for the recommended
reliability indices for use with either the LRFD distribution formulas or refined methodsof analysis such as finite element analysis.
Introduction
For single and multiple-trip special permits that are allowed to mix with traffic(without restrictions on other traffic), the Load and Resistance Factor Rating (LRFR) live
load factors were derived to provide a higher level of reliability consistent with the
AASHTO inventory ratings and Load and Resistance Factor Design (LRFD) design
loading. The prescribed higher target reliability considers the increased risk of structural
damage and benefit/cost associated with very heavy special permit vehicles compared toother truck classes. Though this higher level reliability index is justified based on
structural safety, it has caused operational difficulties for bridge owners because the past
permitting practices have allowed permits to operate at a lower reliability level.
The target reliability index for routine permit crossings is currently established ateither a reliability index beta of 2.5 or 3.5. This reliability index needs to be comparedwith reliability indices used in current routine permit practices and adjusted as
appropriate to meet operational needs.
The live load distribution for special permits is based on the tabulated LRFD onelane distribution factors with the built-in multiple presence factor (i.e., 1.2) divided out.
The live load distribution analysis for routine permits uses LRFD two-lane distribution
factors assuming the simultaneous side-by-side presence of non-permit heavy trucks on
the bridge. The load factors are higher for spans with higher average daily truck traffic(ADTT) and lower for heavier permits. The current LRFR permit load factor calibration
for routine and special permits is tied to the LRFD distribution analysis method and does
not provide guidance on the use of refined methods of analysis for heavy permits or for
permits with non-standard gage widths. Therefore, live load factors and analysis guidance
that are appropriate for analysis methods other than the LRFD distribution formulas need
to be derived.

䋭191䋭

This paper summarizes the research conducted under this project to achieve theresearch objectives. This summary is based on the contractor’s final report authored by
Mr. Bala Sivakumar of HNTB Corp., New York and Dr. Michel Ghosn of the City
College of New York, New York.
Analysis of Representative WIM Data
To verify that the AASHTO LRFR produces acceptable and uniform levels of
reliability for typical U.S. bridges under current loading conditions, it is critical to use themost representative statistical information on truck weights, truck configurations, andmultiple presence data. In this study, weigh-in-motion (WIM) data were analyzed toobtain projections for the maximum bridge load effects from six sites located oninterstate highways in New York, Mississippi, Indiana, Florida, California, and Texas.
The data were gathered in 2005 and 2006 for each traffic direction and included thenumber of axles, axle spacings, and axle weights for each truck. Multiple presence
probabilities were assembled from a representative site in New York (Sivakumar et al2008). Table 1 lists descriptive information for each site including the average daily truck
traffic (ADTT) and the number of truck records after filtering the data to eliminate any
questionable data.
Table 1 WIM Data for LRFR Recalibration
Interstate#Trucks
Site
State
Route
ADTT
Recorded(Direction)

Mean GVW(kips)

Mean of top
10% GVW(kips)

0001

CA

I-5 (E/N)
I-5 (W/S)

1,537,613
1,470,924

5,058
4,839

56.0
53.3

80.6
81.9

0526

TX

I-20 (E/N)
I-20 (W/S)

1,330,799
1,174,954

4,070
3,593

55.6
56.7

80.8
81.5

2606

MS

I-55 (N)I-55 (S)564,393604,919

1,622
1,733

66.5
63.2

108.7
83.5

9121

NY

I-81 (N)I-81 (S)531,042525,733

1,715
1,614

57.2
57.8

101.7
98.3

9512

IN

I-74 (E)
I-74 (W)931,971
1,003,443

2,596
2,795

60.7
60.1

82.5
87.3

9926

FL

I-75 (N)I-75 (S)1,096,076
1,032,680

4,136
3,897

48.4
53.5

84.1
84.6

䋭192䋭

Table 2 Average Reliability Index Values for Random Trucks on 2-Lane Bridges
ADTT=5000
ADTT=1000
ADTT=100
Bridge Type
Moment Shear Moment Shear Moment Shear
T-Beam
1.291
0.978
1.855
1.180
1.855
1.558
Prestressed Concrete
1.083
0.995
1.329
1.204
1.783
1.595
Noncomposite Steel
1.124
0.809
1.361
1.067
1.799
1.547
Composite Steel
1.113
0.793
1.351
1.052
1.790
1.533
Permit Trucks on Two-Lane Bridges
The reliability analysis was performed for the selected bridge configurations
assuming that AASHTO LFR operating rating factor for each permit truck of the set of
typical permit trucks is equal to 1.00. The member resistances that would be required toallow a permit truck to cross each bridge are calculated assuming that the bridges arebeing evaluated using the current AASHTO LFR operating rating criteria with a live loadfactor 1.3 applied on the routine permit and using the two-lane distribution factor.
Table 3 provides a summary of the average values for each bridge type and ADTT
for both moment and shear. The overall average reliability index is 2.94. This averagevalue is considerably higher than that observed for previous case (random trucks onspan). This difference is because the rating process uses the actual permit load andassumes that the random truck alongside the permit is of equal weight. The difference isalso due to the fact that the permit trucks weights are much better known and areassociated with lower coefficient of variation values than the random trucks. Although
the permit truck may still cross the bridge with a random truck, the chances of having apermit truck alongside a truck of equal or higher weight are relatively low. These factors
lead to significantly higher reliability index values for the permit trucks than in the casesof random loading.
Table 3 Average Reliability Index Values for Permit Trucks on 2-Lane Bridges
ADTT=5000
ADTT=1000
ADTT=100
Bridge Type
MomentShear
MomentShear
MomentShear
T-Beam
2.71
2.58
2.75
2.61
2.83
2.68
Prestressed Concrete
2.99
2.66
3.05
2.69
3.25
2.76
Noncomposite Steel
3.01
3.10
2.99
3.14
3.10
3.23
Composite Steel
2.93
3.09
2.98
3.13
3.10
3.23
Permit Trucks on Single Lane Bridges
The reliability analysis approach used for permit trucks on two-lane bridges wasapplied using the one lane distribution factor; the results are presented in Table 4. The

䋭196䋭

average reliability index for all the cases considered is 3.62, which is higher than thatobtained for the permits crossing over two-lane bridges. This difference is because thepermit truck on a 1-lane bridges is alone on the bridge and thus the total applied load isbetter known than in the case where the truck may cross alongside a random truck which
would govern the analysis of permit truck crossing two-lane bridges.
Table 4 Average Reliability Index Values for Permit Trucks on Single Lane Bridges
All ADTTs
Bridge Type
MomentShear
T-Beam
3.60
3.39
Prestressed Concrete
4.01
3.04
Noncomposite Steel
3.91
3.56
Composite Steel
3.91
3.55
Reliability Targets for Permit Load Recalibrations
Target reliability index for the calibration is set at 2.5 with the goal of achievinga minimum reliability index values for all conditions above 1.5.
Permit Load Classifications
The recalibration of the live load factors for permits considered the following fourcases:
I. Permit vehicle alone on a bridge which can occur whether the permit has been
issued for a single trip or multiple trips.
II. Unlimited crossings of multiple trip permits in which two permit trucks could
cross a bridge simultaneously side-by-side.
III. Unlimited crossings where a permit truck mixes with other random vehicles.
IV. Limited number of trips (1 and less than 100) where the permit truck can mix
with other random trucks.
Cases I, II, and III consider routine permits where the actual truck weight may
sometimes exceed the weight limit. Cases I, II, and IV consider special permits where
the weights are assumed to be fully controlled and are not expected to exceed the permit
weight allowed. For routine permits, it is Case III that is expected to govern, while CaseIV should govern for special permits. Case I is analyzed to check the safety of escorted
permit trip. The analysis of Case II is performed to verify that it will be overshadowed
by Cases III and IV.
Case I is not affected by the WIM data for the random trucks. Case II considers
the probability of having two permit trucks side-by-side cross a bridge within the five
year rating period. In the reliability analysis, it assumed up to 100 permits per day as an
upper limit for the number of trips (Moses 2001). The probability of having two side-by-

䋭197䋭

side permits is 0.5% based on the WIM data collected on New York state sites on low
truck traffic volume days (Sivakumar et al, 2008).
For cases III and IV, the reliability analysis should account for the number ofrandom vehicles that may cross the bridge simultaneously with the permit truck.
Following the AASHTO LRFR classifications sites with ADTT of 5000, 1000, and 100
are considered. The percentage of side-by-side vehicles (Psxs) is 2%, 1.25%, and 0.5%
for sites with ADTT of 5000, 1000, and 100, respectively. These Psxs values are upperbounds obtained from the headway data collected at ten WIM sites in New York State(Sivakumar et al 2008).
Key Findings
The calculations performed demonstrate that using a live load factor of 1.10 for
escorted special permit loads will provide average reliability index values greater than 2.5
when the single lane AASHTO LRFD load distribution factors are used to check whether
the permit truck can be allowed to cross a bridge. When performing a refined analysis of
the bridge, it is recommended to use the same live load factor (i.e., 1.10) for escorted
special permits. Special permits travelling over bridges at crawl speed should still bechecked with a dynamic allowance factor of 1.05 to satisfy the minimum reliability index
of 1.5.
For the case when a refined analysis is performed for special permits that may
mix with traffic, the target reliability is also exceeded when a live load factor of 1.00 is
applied on the permit truck while a live load factor of 1.10 is applied on the governing
AASHTO legal truck placed in the adjacent lane.
For the cases of routine permits, where data shows that permit loads may exceed
the permit weight limits, having live load factors varying from 1.40 for sites with ADTTof 5000, 1.35 for sites with ADTT of 1000, to 1.30 for sites with ADTT of 100 will leadto average reliability index greater than 2.5 while the minimum reliability index values
remain above 1.5. These checks should be performed with the two-lane AASHTO LRFD
load distribution factors.
The above live load factors for routine permits can be reduced for the cases where
the permit truck’s gross vehicle weight is high in order to reflect the lower probability of
having a random truck of equal or higher weight crossing alongside the permit truck.
Specifically, for trucks with GVW/AL < 2.0 kip/ft (i.e., gross vehicle weight (GVW) overfront axle to rear axle length (AL)), it recommended to use the above mentioned live loadfactors (1.40, 1.35, and 1.30 for sites with ADTT of 5000, 1000, and 100, respectively).
For trucks with GVW/AL between 2.0 and 3.0 kip/ft, the recommended live load factors
are 1.35, 1.25, and 1.20 for sites with ADTT of 5000, 1000, and 100, respectively. For
trucks with GVW/AL above 3.0 kip/ft, the live load factors are 1.30 1.20, and 1.15 forsites with ADTT of 5000, 1000, and 100, respectively.

䋭198䋭

The research also recommended revisions to LRFR permit rating specifications
with commentary suitable for inclusion in the AASHTO Manual for Bridge Evaluation
including a new table of LRFR permit load factors.
Acknowledgments
The work summarized herein was performed under NCHRP Project 20-07, Task
285 and was guided by NCHRP Project Panel 20-07, Task 285. Dr. Waseem Dekelbab
served as the responsible NCHRP staff officer. The final report was prepared by Mr. Bala
Sivakumar of HNTB Corp., New York and Dr. Michel Ghosn of The City College of
New York, New York.
References
Moses, F. (2001). “Calibration of Load Factors for LRFR Bridge Evaluation,” NCHRP
Report 454, Transportation Research Board, National Research Council, Washington,
DC.
Sivakumar, B., Ghosn, M., and Moses, F. (2008). “Protocols for Collecting and Using
Traffic Data in Bridge Design,” Unpublished draft final report for NCHRP Project 12-76,
National Cooperative Highway Research Program, Transportation Research Board,
Washington, DC.
Sivakumar, B. and Ghosn, M. (2011). “Recalibration of the LRFR Load Factors in the
AASHTO Manual for Bridge Evaluation,” Final report for NCHRP Project 20-07/Task
285, National Cooperative Highway Research Program, Transportation Research Board,
Washington, DC.

䋭199䋭

䋭200䋭

COMPRESSIVE LOADING TEST OF CORRODED GUSSET PLATECONNECTION IN STEEL TRUSS BRIDGE
Jun Murakoshi1, Naoki Toyama1, Mamoru Sawada1, Kentaro Arimura1, Lu Guo1
Kuniei Nogami2, Teruhiko Yoda3, Hideyuki Kasano3
Abstract
With the stock aging of the majority of highway bridges in Japan constructedduring the 1950s–1970s, some serious corrosion deterioration cases of fracture criticalmembers in steel truss bridges have been reported recently. In this paper, compressive
loading test of severely-corroded gusset plate connections cut out from a demolishedtruss bridge were conducted in order to assess the remaining load capacity.
Introduction
The majority of highway bridges in Japan were constructed during the1950s–1970s which coincides with Japan’s high economic growth period, and thenumber of bridges over 50 years is increasing drastically. With increase of aged bridges,
since these bridges are exposed to heavy traffic and severe natural environment, it ishighly probable that the deterioration and damage will increase rapidly. Improvementof technologies related to inspection, diagnosis, repair, and rehabilitation needed.
Concerning steel bridges, some serious deterioration cases of FCMs on steel truss
bridges have been reported recently. A tension diagonal member of steel truss
embedded inside the deck concrete fractured in the Kiso River Bridge and Honjo
Bridge on the National Route because of corrosion that invisibly progressed inside theconcrete in 2007. Fracture of diagonal members or gusset plate connections of truss
bridge is likely to lead to fatal damage of whole bridge. On the other hands, there wasno effective measure to evaluate remaining strength of such deteriorated components
and the whole bridge system with the uncertain section loss from corrosion.
The authors initiated research project in order to identify the remaining load
capacity and to investigate how to evaluate the remaining strength of deteriorated
diagonal members and riveted gusset plate connections subjected to severe corrosion.
In this research project, several corroded specimens are going to be tested within a few
years. These specimens consist of diagonals and gusset plate connections which were
cut out from demolished steel bridges which were in service about 50 years nearcoastal area.
This paper reports the preliminary results from of compressive loading test of
the first one specimen conducted in September, 2011, and discusses compressive
behavior and the ultimate strength for severely-corroded gusset plate connection.
1

Center for Advanced Engineering Structural Assessment and Research(CAESAR),
Public Works Research Institute(PWRI)
2
Department of Civil and Environmental Engineering, Tokyo Metropolitan University
3
Department of Civil and Environmental Engineering, Waseda University

䋭201䋭

Before the test, section loss was measured using laser measurement equipment, and theeffect of the section loss on failure behavior and the ultimate strength were examined
by Finite Element analyses to complement experimental results. Then the authors
compare experimental results with analysis results and strength equations in gusset
plate connections.
Bridge Description
Figure 1 shows a bridge utilized in this project, which is called Choshi Bridge.
It was built in 1962 across Tone River, called Choshi Bridge. It was 5-span steel
through truss bridge with total length of 407.4m. Figure 2 shows general and section
view of the bridge. The average daily traffic is about 20,000 with 10% of heavy
vehicles. It was located in river mouth and had suffered from salt damage by airborne
salt and heavily corroded. Although repainting, strengthening and partial replacementof severely corroded members were conducted several times through its service life, it
was finally replaced in 2009 at 47 years old, because the corrosion was unlikely to stopand it is considered to be impossible to assess remaining strength and remaining
service life.
Figure 3 shows corrosion damage focusing on main members and gusset plateconnection that influence safety of the whole bridge. Steel members of this bridge havebeen repainted by the thick fluorine coating material, so section loss was not able to beobserved exactly by visual inspection. Corrosion of gusset plate connections are shownin Figure 3(a) (b). Several connections and diagonals were strengthened with steel
plate bonding (see Figure 3(c) ). Intense corrosion of diagonal joint is shown in Figure3(d). Pitting of diagonal was observed in Figure 3(e). Concerning floor beams, Figure
3 (f) shows typical area of deterioration of floor beam with debris accumulation.
Compression Load Test
Specimen Description and Experimental Setup
After demolished, several connection parts and diagonal members were cut out
as experimental specimens, and carried to our laboratory after the coating was removed.
For the present, we are planning to conduct the loading test for 4 specimens which havedifferent gusset configurations. Figure 4 shows the first one test specimen, which was
cut out from upper chord connection P25d near intermediate support. The diagonal issquare box type section with flange of 500mm width and 10 and 12 mm thickness at theconnection, and thickness of gusset plate is 12mm. Design axial force/stress of the
diagonal members are listed in Table 1. Steel grade is SM40 (400MPa nominal tensile
strength), the yield strength is 284MPa by tensile material test of the diagonal member.
Section loss at the outer and inner surface of the specimen was measured using
laser surface measurement equipment (see Figure 5). The measurement interval was set
to 1mm to understand the mechanical behavior for uneven surface. As it was difficult
to measure the inner surface directly, the surface shape was taken using plaster, andthen it was measured. Figure 6 shows contours of corrosion areas. Red area means

䋭202䋭

large section loss, and yellow color means non-corrosion areas. Severe section loss wasobserved at connection parts of diagonal and gusset plate. As for the gusset plates,
severe sections loss on the outer surface was not be seen except the rivets areas. Severe
section was observed on the inner surface, where humidity seems high and airborne salt
is likely to accumulate. As for the compression diagonal, large section loss on the outersurface was hardly found except the edge of flange, however, large section loss wasshown on the inner surface around the gusset plate boundary. The maximum corrosion
depth on the inner surface of the compression diagonal is 8.0mm (thickness of the
diagonal flange: 12mm), the average corrosion depth is 3.4mm. The maximum and theaverage corrosion depths on the inner surface of the gusset plate are 9.0mm and 4.0mm
respectively. The average corrosion depth at the plate area underneath the diagonal is
6.7mm. The average remaining thickness of the gusset plate is 8.0mm. The average
reduction area ratio is 19% for the compression diagonal and 33% for the gusset plate.
Comparing the measured section loss distribution with FE analysis results, it was
found that severe corrosion part generally corresponded to the part where large stressappears. As a result, gusset plate connections may be structural weakpoint.
Figure 7 shows outline of specimen and loading frame. Figure 8 shows
experimental setup. The compression and the tension axial loads were applied to the
diagonal members at the same load increment step, because the absolute values of thedesign axial forces of both diagonals are almost equal. However, by the restrictioin of
capacity of tension jack, tensile load was fixed to 2000kN. 30MN testing machine forcompression and loading frame with jacks for tension were used for bi-axial loading.
Analysis Method
FE analyses were carried out to investigate the effect of section loss oncompressive behavior by using a model shown in Figure 9. The analysis model
simulated test condition. In modeling, 4 nodes shell elements were used for gusset
plates and diagonals. Rivet fasteners were modeled by spring elements. The
stress-strain relation of steel was assumed to be bi-linear, with a second modulus of
E/100(E=2×105MPa). Upper chord was restrained with the loading frame atconnection part. The displacement along the loading direction at the loading point isfree, and the displacements of two other directions are fixed. In this analysis, the initial
imperfection is not considered.
Analyses were conducted for two cases of non-corroded and corroded model
simulating test specimen. Figure 10 shows assumed plate thickness of corroded model
which reflects the measured data. Average thickness reductions were 2.0mm for the
diagonal flange, 3.0mm for the diagonal web and 4.0mm for the gusset plate,respectively.
Experimental and Analysis Results
Figure 11 shows the curves of load versus vertical displacement at the loading
head. The analytical ultimate strengths were 4953kN for the un-corroded model and3346kN for the corroded model. The ratio of the strengths is about 2/3, which is similar䋭203䋭to average thickness loss of the gusset plate. The measured ultimate strength was
3598kN, that is about 1.1 times the analytical value for the corroded model. Linear
behavior was observed until the out-of-plane deformation of gusset plate become large.
After that, the load reached maximum load gradually and fell down moderately. The
measured value and analytical value show generally the same curves and ultimate
loads.
Figure 12 shows failed specimen after the test. The failure mode of thespecimens was plate local buckling of the gusset. Figure 13 shows out-of-plane
deformation and relations between load and the deformation of the both side of gusset
plates at major points. With increase of vertical load, deformation of one side of thegusset plate preceded with the other side of the gusset. As a result, the buckling shape of
unsupported edge shows unsymmetry. As for the analytical results of the corroded
model, Von Mises stress contours and yielded area at the peak load are shown in Figure
14 and Figure 15, respectively. The local buckling occurred at the plate area
underneath the diagonals and free edges of the gusset plate. Figure 16 compares theout-of-plane deformation at major points where large deformations were measured andshows good agreement. For reference, analytical out-of-plane deformation contours of
the corroded model are also shown in this figure. The results in these figures provide
verification of the corroded model using shell element to evaluate compressive
behavior of the corroded gusset plate connection. About the modeling of the corrosion,
the use of average reduction thickness of gusset plate seems reasonable to evaluate thebehavior of the gusset plate in this specimen, however detailed investigation is required.
Figure 17 shows the out-of-plane displacement along the line parallel to the centerlineof the compression diagonal.
Strength Estimation Equations of Truss Gusset Plate Connections
Strength Equations
After the collapse of I-35W Bridge, “Load Rating Guidance and Examples for
Bolted and Riveted Gusset Plates in Truss Bridges” [2] was issued by FHWA in 2009.
By referencing the Guidance and previous experimental research results [3]- [7], limit
state of gusset plate and diagonal members are assumed as follows as shown in Figure
18,a) Strength of fasteners in compression and tensionb) Cross section yielding or net section fracture strength of gusset platec) Block shear rupture strength in tensiond) Cross section yielding or net section fracture strength of diagonal membere) Compressive strengthf) Shear fracture strength
This paper only discusses compressive strengths of b), d) and e). The resistancefactors are1.0 in this study.

䋭204䋭

Cross section yielding strength of gusset plate in compression
The Whitmore effective width[3] is used for estimating yielding of the gusset
plate. The effective width is bound on either side by the closer of the nearest adjacentplate edges or lines constructed starting from the external fasteners within the first row
and extending from these fasteners at an angle of 30 degrees with respect to the line ofaction of the axial force (see Figure 19). The cross section yielding is taken as:
Pgy

(1)

f y Aewhere:
Ae:gross cross-sectional area of Whitmore effective width of the plate, Ae=Let(mm2)
fy: yield strength of the plate (N/mm2)
Le:Whitmore effective width (see Figure 19)(mm)
t: thickness of the plate (mm)Cross section yielding of diagonal member
The smallest sectional area of the diagonal members near the gusset plateboundary is assumed to be yielded. The cross section yielding strength is expressed by:Pdy

f y Ag(2)where:
fy: yield strength of the diagonal (N/mm2)
Ag: gross cross-sectional area of the diagonal (mm2)Local buckling at the plate area underneath the splice member of diagonals
The Whitmore effective width and an unbraced gusset plate length which isaverage of the three lengths was used for estimating buckling strength. Standard
buckling equations specified in Japanese Design code (JSHB) was used. Ignoring any
lateral constraint to the gusset plate, the effective length factor, β (β=1.2) was used for
unbraced gusset plate assuming the buckled shape as shown in Figure 20. The local
buckling equation is taken as:
Pgcr―(Ȝ҇0.2)f y Ag

Pgcr (1.109  0.545O ) f y Ag
Pgcr (1.0 /(0.773  O ) f y Ag
2where:
fy: yield strength of the plates (N/mm2)
Ag: gross cross-sectional area (mm2)

䋭205䋭

―

(0.2 <Ȝ҇1.0)―(1.0<Ȝ)

(3a)
(3b)
(3c)

̆

The column slenderness ratio ǳ is given by:O

f y EL
1
࡮ ࡮ c
E rs
S

(4)

where:
E: Young’s modulus of plate (N/mm2)
ȕ: effective length factor (=1.2)
Lc: Lc = (L1+L2+L3)/3
L1, L2, L3: distance from center or each end of the Whitmore width to the edge in theclosest adjacent member, measured parallel to the line of action of the compressive
axial force (see Figure 19).
rs: radius of gyration about the plane of buckling, rs
I g / Ag (mm)Ig : moment of inertia (mm4)Comparison of Analysis Results and Calculation Results
Table 2 outlines the comparison of the experimental results, FE analysis results
and the calculation results for the specimen. The ratio means the calculated or
measured value to the analytical value. The calculated yield strength by the Whitmore
effective width was to some extent close to the analytical ultimate strength with ratios
of 0.97 (un-corroded model) and 0.95 (corroded model). On the other hand, thecalculated yield strength of the diagonal was larger than the analytical value with ratiosof 1.23 and 1.39. It is indicated that the gusset plate failure preceded with yielding of
the diagonal. Strength equation for local buckling gives conservative estimates withstrength ratio of 0.59 (un-corroded model) and 0.36 (corroded model), much below 1.0.

Regarding the compressive strength of the gusset plate connection, the resultsin this study were compared with experimental results[4]-[8]. Figure 21 shows
comparison of the measured ultimate loads and the calculated values for local bucklingand yielding respectively. Figure 22 shows relations of ultimate strength andslenderness ratio. Calculated values are also conservative for the experimental data,
and the correlation is not good. Then, we are investigating more accurate estimation ofultimate strength of the gusset plate. According to the failure mode, the ultimate
strength is likely to depend on the buckling strength of the compressive unbraced area
parts and the strength of its surrounding plate area. As one of our ideas, we are trying
to evaluate the compressive strength by the summation of following strength equationsof gusset plate divided into 3 areas as shown in Figure 23.
Pgcr

Pgcr1  Pgcr 2  Pgsy

䋭206䋭

(5)

Pgcr1 is expressed by:

―

Pgcr1
Pgcr1

(Ȝ҇1.0)f y Ag
1

O

2

(6a)

―

f y Ag(1.0<Ȝ)(6b)

̆

The column slenderness ratio ǳ is given by:O

f y EL
1
࡮ ࡮ c
E rs
S

(7)

where:
ȕ: effective length factor (=0.65)
Lc: Lc = (L1+L2+L3) / 3
L1, L2, L3: The distance from center or each end of the width of diagonal end to theedge in the closest adjacent member, measured parallel to the line of action of thecompressive axial force (see Figure 23).
Pgcr2 is expressed by:
Pgcr 2f y Ag sin T1(R҇1.0)(8a)

Pgcr 2

1
f y Ag sin T1
R2(1.0<R)(8b)

The plate slenderness ratio R is given by:Rb f y 12(1  Q 2 )
࡮ ࡮
t
E
S 2k

where:
ǵ: The Poisson's ratio (=0.3)
k: The buckling coefficient , k

ǩ: ǩ=hc / b2

4

D


2

(9)

40 15D 2 20Q
 4  2SS
3S 2

hc: hc=(h1+h2) / 2
Pgsy is expressed by:
Pgsy

fy
3

Ag cosT 2(10)

Figure 24 shows comparison of the measured ultimate loads and the calculated
values. It is noticed that failure modes of all data are local buckling, not compressive
and block shear failure which is described in [8]. Considering that previous
experimental data contain various gusset configurations, it appears the ultimate
strength can be approximately estimated. Still there is a difference, further study isrequired to estimate the ultimate strength for compressive load.

䋭207䋭

Conclusions
Compressive loading test of the corroded gusset plate connection specimen
from decommissioned truss bridge was performed, and the FE analyses wereconducted to complement experimental results. As for compressive strength estimationof gusset plate connection, from practical viewpoint, application of strength equations
were discussed with use of previous experimental research results. The major findings
are summarized as follows.
1) Based on thickness loss measurement of gusset plate connection, advanced
corrosion of diagonals and gusset plate was observed around the connection
parts. Severe corrosion part generally corresponded to the part where large
stresses appear.
2) The effect of the section loss on the compressive strength of the gusset plate was
evaluated by experimental and analytical results. Compressive behavior of thegusset plate was properly evaluated by shell element model in consideration of
the average thickness reduction.
3) Local buckling strengths by the Whitmore effective width provided conservative
estimates to the experimental ultimate strength. Taking the buckling strength ofthe compressive area and the strength of its surrounding plate area into
consideration gave more proper prediction.
Acknowledgment
This research was undertaken as part of the collaborative research project
between Public Works Research Institute; Tokyo Metropolitan University; andWaseda University, and funded by the Ministry of Land, Infrastructure, Transport andTourism based on the Construction Technology Research and Development Subsidy
Program. Finally, the authors express appreciation to Choshi Public Works Office,Chiba Prefecture for their cooperation.
References[1] Japan Road Association (JRA), “Specification for Highway Bridges, Part II Steel
Bridge”, 2002.(in Japanese)[2] Federal Highway Administration, “Load Rating Guidance and Examples For
Bolted and Riveted Gusset Plates In Truss Bridges”, Publication
No.FHWA-IF-09-014, 2009.
[3] Whitmore, R.E., “Experimental Investigation of Stresses in Gusset Plates, Bulletin
No.16, Engineering Experiment Station”, University of Tennessee,1952.
[4] Yam, M. and Cheng, J., “Experimental Investigation of the Compressive Behaviorof Gusset Plate Connections”, Structural Engineering Report No.194, Dept. of
Civil Engineering, University of Alberta,1993.

䋭208䋭

[5] Ocel, J. M., Hartman, J.L., Zobel, R.,White, D. and Leon, R., “ Inspection and
Rating of Gusset Plates - A Response to the I-35W Bridge Collapse”, Proceedingsof the 26th US-Japan Bridge Engineering Workshop, pp.11-23, 2010.
[6] Matsuhisa, S., Yamamoto, K. and Okumura, T., “Loading Experiment of Truss
Gusset Plate Connections”, Proceedings of the 31st Annual Conference of Japan
Society of Civil Engineers, pp.297-298, 1976. (in Japanese)[7] Matsuhisa, S., Yamamoto, K. and Okumura, T., “Loading Experiment of Truss
Gusset Plate Connections”, Proceedings of the 32nd Annual Conference of Japan
Society of Civil Engineers, pp.631-632, 1978.(in Japanese)[8] Kasano, H., Yoda, T., Nogami, K., Murakoshi, J., Toyama, N., Sawada, M.,
Arimura, K. and Guo, L., “Study on Failure modes of Steel Truss Bridge Gusset
Plates Related to Compression and Shear Block Failure”, Proceedings of the 66th
Annual Conference of Japan Society of Civil Engineers, pp.149-150, 2011. (inJapanese)䋭209䋭27th US-Japan
Bridge Engineering Workshop

Session6

Maintenance2
Introduction of Non-Destructive Highway Inspection Methods Using High Definition Video andInfrared Technology
By Koji Mitani and Masato MatsumotoCharpy Impact Tests with Test Specimens Made with Stop-Hole-Size Cores
By Kiyoshi Ono, Kengo Anami, and Mitsuharu Oikawa
The Next Generation of Experimental Research in Structural Fires
By John L. Gross and Stephen A. Cauffman
New Attempt to Maintenance of Steel Bridge Coating in Expressway
By Shuhei Sakai and Dai Wakabayashi
FHWA’s New Infrastructure Research and Technology Strategic PlanBy Ian M. Friedland

Page Left Blank

䋭220䋭

INTRODUCTION OF NON-DESTRUCTIVE HIGHWAY INSPECTIONMETHODS USING HIGH DEFINITION VIDEO AND INFRAREDTECHNOLOGY
Koji Mitani1 and Masato Matsumoto2
Abstract
In order to make timely and appropriate maintenance and rehabilitation
decisions for deteriorating highway structures, constant monitoring of structural
conditions is necessary. While most agencies recognize the importance of timely
bridge inspections, such programs tend to be time consuming and expensive. Under
these circumstances, NEXCO-West has been able to reduce highway structure
inspection costs by introducing an innovative highway inspection technical approachthat uses a combination of high definition video (HDV) and infrared (IR)thermographic technology. This paper describes these inspection methods using HDV
and infrared technologies and introduces some examples of practical on-site
application to highway bridge superstructures.
1. Introduction
Today, proper maintenance and management of deteriorating infrastructure
under severe budget constraints have become serious issues for bridge owners.
Traditionally, highway bridge conditions have been monitored by visual inspection
with structural deficiencies being manually identified and classified by qualified
engineers and inspectors. However, the quality of inspection results obtained throughthe traditional approach depends on the individual inspector’s subjective judgment
based on his/her knowledge and experience. In addition, these traditional inspection
procedures require significant investments in both time and labour cost. These factors
support the necessity for research and development for more reliable, objective andefficient bridge inspection methods.
With traditional site inspections, qualified inspectors are performing close-up
visual inspections and sounding tests, often from crane suspended lifting cages orbuilt-in inspection staging; arguably putting inspectors at some safety risk. The needfor safer inspection methods calls for new innovations in bridge inspection
technologies. In addition, new technologies that improve inspection efficiencies will
help address the upcoming shortage of qualified bridge inspectors.
Bridge inspector responsibilities include preparing summaries of bridge
condition factors that, by their nature, reflect the individual inspector’s engineering
judgement. If we can improve data collection efficiencies and reduce the time required
1
2

President and CEO, NEXCO-West USA, Inc.
Executive Vice President, NEXCO-West USA, Inc.

䋭221䋭

by inspectors in the field to make general structure condition assessments, more time
will be made available for these same inspectors to perform detailed hands-on
inspection for those pre-screened bridge elements where structural defects require
special attention.
The West Nippon Expressway Company Ltd. (NEXCO-West) has beenworking to develop efficient non-destructive highway bridge inspection methods using
High Definition Video (HDV) and Infrared (IR) Imagery technologies. This paper
describes the mechanisms of these inspection technologies and presents results from an
on-site pilot project performed to evaluate the feasibility of these technologies in
Florida, USA.
2. New Inspection Technologies
2.1 Inspection Methods Using High Definition Video 㸦HDV㸧
Recently, research and development on crack detection methodologies for
efficient highway bridge inspection using digital images of the structures have seen
significant technological progress. In the past, conventional inspection techniques
using digital image processing had not been widely applied for practical use due to its
limited image quality. The equipment was typically expensive and their application
was limited primarily to technical research applications and special forensic
professional services. However, recent innovations and improvements in image qualityand data processing technology have contributed greatly to the technical viability ofthis inspection technology.
FIGURE 1 shows the mechanism of pavement crack detection using HDV. The
technology is the combination of GPS, GIS and HDV image pictures. The GPS
navigation system, HDV and laptop computer are included inside the inspection
vehicle. The HDV camera is attached on top of the inspection vehicle to record thesurface condition (FIGURE 2). The recorded data is analyzed by image processing todetermine an individual structure’s current condition as related to crack size, locationand distribution. The detected cracks are identified in a digital crack map. The crack
size and length are determined by computer software, and these quantitative
characteristics are also summarized in spreadsheet format. The obtained crack maps
and related data are provided to engineers for their subsequent structural diagnosis andrehabilitation planning.
A special advantage of HDV technology, with respect to crack identificationand measurement, is the ease of maintaining a historical record of bridge cracks for use
in monitoring crack propagation over time. The image processing includes a twogradation analysis and line featuring analysis. The first step of the two-gradation
analysis converts the digital picture into binary (black-and-white) data by analyzing thedegree of color transition from the nearby pixels, enabling our computer program todifferentiate the spectrum and identify individual cracks. The second step thenidentifies the lines of black pixels in order to confirm the existence of cracks (seeFIGURE 3).

䋭222䋭

Keys (see FIGURE 7). Currently, condition of the bridge is regularly monitored
through established visual inspection procedures performed by qualified inspectors.
3.2 The Pilot Project Results(1) Deck Surface Inspection Using HDV
In order to record the deck surface cracks using an HDV camera, proper height(approximately 3 meters) and recording angle (no greater than 45° from vertical) are
required. The HDV camera was attached to a custom-made camera mount and video
data was gathered facing in the backward direction. HDV recordings of the concrete
deck surface were conducted at a speed of 70km/h.
FIGURE 8 shows an example of a crack map for a concrete deck surface.
Cracks of 3mm or greater were detected by a software supported automatic crack
detection program, followed by supplemental manual crack checking by anexperienced engineer. Manual crack checking successfully detected cracks of 0.8mm
or greater. According to the Bridge Inspectors Field Guide (Florida Department ofTransportation (2008)), cracks should be classified into three categories as shown in
Table 2, and the NBI (National Bridge Inventory) specified “Distressed Area” iscalculated for the rectangular area including “Significant,” “Moderate,” or “Severe”
cracks. Inspectors are responsible for proposing priorities on rehabilitation to thebridge owners by comparing the “Distressed Area” for each span or bridge. The resultsof pilot area bridge deck surface inspections proved the accuracy of crack detection
using HDV technology to be satisfactory for routine in-service deck inspections. The
new inspection technology provides additional benefits by increasing the level ofsafety for both inspectors and motorists and storing position recorded historicalinspection data for monitoring of crack propagation. The digital crack map databasecan be a powerful tool for supporting those engineers responsible for maintenance plan
preparation and work task priority decision-making.

FIGURE 7 - LOCATION OF THE SEVEN MILE BRIDGE

䋭226䋭

that have been covered with fiber reinforced polymer (FRP) materials (see FIGURE
13). Another application is to identify subsurface irregularities (grout covered rockpockets) in new construction structure immediately after the removal of form work.
)53 LQ 3ODFH DIWHU &RQFUHWH 3DWFKLQJ,QIUDUHG ,PDJH'DPDJH &ODVVLILFDWLRQFIGURE 13 DETECTING POTENTIAL SPALL BENEATH THE FRP COVER

5. Summary and Conclusions
This paper described the mechanism of non-destructive bridge inspection
methods using HDV and IR imagery technology and results of the on-site pilot project
performed to evaluate the feasibility of applying these technologies for in service
bridge inspection in the State of Florida, USA.
It was verified from the pilot project results that the accuracy of detection andmeasurement surface cracks and potential subsurface deterioration using these newtechnologies provided satisfactory and acceptable results for practical routine andspecial condition bridge inspections in compliance with recognized inspection
practices. It was also demonstrated that new HDV and IR technologies could
significantly reduce site inspection times and on-site inspection resource requirements.
With the quantity of roadway structure assets increasing annually, coupled withconcurrent increasing rates of deterioration being experienced by many of the existing
structures, bridge owners need to find new and creative ways to ensure the structuralsafety of their bridges while they all too often face problems of reduced budgets anddedicated bridge inspection resources. Using the proposed new HDV and IR
technologies, bridge engineers can quickly and efficiently obtain objective current
bridge condition information that has traditionally been obtained by more time
consuming and, in some instances, more subjective close-up visual inspections andsounding tests. The digital output of these HDV and IR inspection techniques improves
on-site inspection safety and objectivity and contributes to improved inspector
efficiency by reducing significantly the amount of on-site inspection time in the field.
However, it must be noted that while HDV and IR technologies do offer new
efficiencies to the bridge inspection process, they are not a substitute for inspectors
conducting on-site specific follow-up and detailed structure investigations. While䋭230䋭

improved efficiencies in bridge inspection brought about by the application of HDV
and IR technologies bring significant benefits to the overall bridge inspection process,they are not a substitute for the continued need for sound experienced engineering
judgement.
Currently, costs of traditional and new inspection technologies are similar withnew tech's lower field data collection costs being somewhat offset by additional costs
for computer supported analysis. We expect with likely improvements in computer
technology that these new inspection technologies will become increasingly costeffective. The authors believe that by offering experienced bridge engineers andinspectors new improved inspection technologies, bridge inspection programs will bestrengthened through improved inspection data, increased safety and more economical
operations…bringing tangible benefits to bridge owners and the motoring public alike.

Acknowledgments
The authors would like to express their sincere gratitude to Florida Departmentof Transportation, Transfield Services, Inc., The Monroe County Sheriff's Department,
The Louis Berger Group Inc., Berger Avart Inc., Mr. Per Fostvedt of Infrared Services,
Mr. Harry Stoltzfus of Harcon Bridge Access Services, Dr. Masanobu Sugimoto of
Fuji Engineering Co. Ltd., Mr. Kazuaki Hashimoto of NEXCO-West Engineering
Shikoku Co. Ltd., and Mr. Yuji Uchida of Ergo Vision Co. Ltd. for their cooperationand support during the pilot project.

References
1. Akashi K, Hashimoto K, Koide H, and Togawa M (2006.9): Development ofInspection Method using Infrared Thermography Technology and its Technical
Consideration, Proceeding of JSCE 61st Annual Meeting, PP 1113-1114
2. Florida Department of Transportation (2008.2): Bridge Inspectors Field Guide
Structural Elements,

䋭231䋭

䋭232䋭

CHARPY IMPACT TESTS WITH TEST SPECIMENS MADE
WITH STOP-HOLE-SIZE CORES
Kiyoshi Ono1, Kengo Anami2 and Mitsuharu Oikawa3
Abstract
In order to examine or repair fatigue damaged steel bridges, it is necessary toobtain information of mechanical and chemical properties of steel of damaged members
or joints. However, for the old bridges, it is sometimes difficult to obtain such kinds ofinformation from design articles. For such case, a sample material might be taken fromthe structures, but the sample should be as small as possible.
This study examines the use of small steel pieces regarded as cores, which aretaken from stop-holes or bolt-holes. Test specimens for Charpy impact test are madewith small steel pieces by Electron Beam Weld (EBW) and the effect of the steel piecesize on Charpy absorbed energy are examined.
Introduction
Fatigue damage in steel bridges have been reported (Japan Road Association.
2007, 2009; Miki et al. 2007)and the number of reported fatigue damage has increased.
Some of the fatigue damage are serious fatigue damage such as the fracture that
extended halfway through the circumference of a steel pipe column of a pedestrian
bridge and the crack on the web plate of a steel girder whose length is about 1.1m. In
order to identify the cause of fatigue damage or examine the retrofitting methods offatigue damage, it is necessary to gain the information on the mechanical, chemical andfracture properties of materials of steel bridges. Especially as for the old steel bridges,
there are some cases that the information about the material does not remain or the
standard of the material does not exit at all. In such cases , the test specimens or samples
are picked from the base material of steel bridges. For example, test specimens ofCharpy impact tests were made with the material removed form the base material andCharpy impact tests were carried out in order to examine the fracture toughness (Mikiet al. 2009). However, it is desirable that the area picked from the sound base material
is as small as possible.
By the way, there are cases that stop-holes are bored for preventing the crack
propagation and the holes for high tension bolts are bored for retrofitting by the bolted
splice steel plates. It is very effective and useful to use the stop-hole-cores orbolt-hole-cores as the test specimens or samples.

1

Associate Professor, Dept. of Civil Engineering, Osaka University
Associate Professor, Dept. of Civil Engineering, Shibaura Institute of Technology
3
Bachelor, Dept. of Civil Engineering, Shibaura Institute of Technology
2

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In order to evaluate the fracture toughness by using stop-hole-size cores, thisstudy examines to make the test specimens with the stop-hole-size cores and conduct
Charpy impact tests by using the test specimens. To be concrete, a small steel piece
including the V-notch area gained from the stop-holes or bolt-holes is connected to theother parts of the test specimen by electron beam weld (EBW) as shown in Figure1. The
effect of the width of the small piece on the Charpy absorbed energy is investigated.
Mechanical and chemical properties of material
One of the major interest of this study is to investigate the material properties ofthe old steel. Therefore, the steel plates for the test specimens were picked from a web
plate at a cross beam of a steel bridge constructed in the 1920's or 1930's and test
specimens were made with the steel plates. The plate thickness of the targeted steel plateis 9mm. The information on the material properties does not remain at all. Therefore, thetensile tests and the chemical analysis were carried out in order to get the basic
information on the material properties before Charpy impact tests.
Test specimens were removed in longitudinal direction of the cross beam anddirection perpendicular to longitudinal direction. The number of test specimens in eachdirection is three and the total number is six. Table 1 indicates not only the averagevalues of major mechanical properties about the strength and elongation in eachdirection but also the values specified in the 2008 JIS (Japanese Industrial Standards) as
for SS400 that is rolled steel for general structures and SM400A that is rolled steel for
welded structures. Figure 2 shows an example of stress-strain relationship gained fromthe tensile tests. As shown in Table 1, the mechanical properties of the steel satisfy the
specifications of SS 400 and SM400A in the 2008 JIS.
Table 2 shows chemical analysis results and chemical components specified in
the 2008 JIS in respect of SS400 and SM400A. As shown in Table 1, although theamount of "S" does not satisfy the specification as for SM400A in JIS, the othercomponents in Table 2 satisfy the specifications as for SS400 and SM400A in JIS.
Assumed size of cores
The main purpose of this study to make the test specimens with thestop-hole-size cores for Charpy impact tests in order to evaluate the fracture toughness.
Figure 1 shows an outline of test specimen examined in this study. As shown in Figure
1, a small steel piece including the V-notch area corresponding to a stop-hole core or abolt-hole core is connected to the other parts of the test specimen by EBW. It is assumed
that a diameter of the stop-hole or bolt-hole is 24.5mm or 26.5mm. It is thought that thestop-hole-size core whose diameter is 20mm can be obtained from the assumed
stop-hole or bolt-hole. Figure 3 shows the relationship between an assumed
stop-hole-size core gained from steel bridges and a small steel piece including the
V-notch area of a test specimen. As shown in Figure 3, the square steel pieces whose
size is 13mm can be gained from the stop-hole-size core whose diameter is 20mm. In

䋭234䋭

consideration of Figure 3, 13mm is decided as the maxim size "B" of square steel pieces
gained from the stop-hole-size cores.
Procedure of making test specimens by EBW
The welding condition of EBW used in this study is as follows.
Voltage: 60(kV), Current: 65(mA), Welding speed: 650mm/min.
EBW was carried out from the one side of test specimens. The outline of theprocedure of making test specimens is as follows.
࣭Small steel pieces corresponding the main part of test specimens including V-notch(ձ in Figure 4), steel spacers (ղ in Figure 4), edge parts of the test specimens (ճin Figure 4) and the bucking metal are set as shown in Figure 4. The steel grade ofedge parts of the test specimens is SM490.
࣭EBW (Blue solid lines in Figure 4) is conducted.
࣭Cutting along the yellow dotted lines in Figure 4 is conducted.
࣭ Making test specimens are completed by cutting and shaving off until thepredetermined size and thickness of under-size Charpy V-notch impact test
specimens specified in JIS.
Types of test specimens
It was impossible to make the standard-size Charpy V-notch impact test
specimens whose thickness is 10mm because the thickness of targeted steel plates is
9mm. Therefore, under-sized Charpy impact V-notch test specimens whose thickness is
7.5mm were made in this study. In the case of making test specimens by welding, theinfluence of welding on the change in the material properties like the heat-affected zone
should be considered. Moreover, some previous studies (Seo et al. 1982; Seo et al.
1983) pointed out that Charpy absorbed energy is effected by restriction by EBW andthe adequate test results can not be gained if the distance between each EBW is small.
Therefore, test specimens were set with focusing on the width of small steel pieces "B"in Figure 5. "B" corresponds to the distance of the center of each EBW as shown in
Figure 5. Table 3 shows types of test specimens and the values of "B". In Table 3, thetest specimen "B-0" indicates the test specimens without EBW. According to theexperience of EBW until now, it is supposed that the width of the heat-affected zone by
EBW whose welding condition is almost the same as that in this study may be 6mm. The
width of the non heat-affected zone of each test specimens is supposed to be (B-6)mm
as shown in Figure 5. Judging from this assumption about the width of the heat-affectedzone by EBW, it is estimated that all over the V-notch area of test specimens "B-4" maybecome the heat-affected zone by EBW.
Macrostructure tests and Vickers hardness tests
Macrostructure tests and Vickers hardness tests were conducted in order toinvestigate the heat-affected zone by EBW. The weight of Vickers hardness test in thisinvestigation was 9.8N. The location of the measuring points is the center of the height䋭235䋭

of test specimens and the interval of the measuring points is 0.5mm. Pictures 1 and 2
show the results of the macrostructure tests of the test specimens "B-1" and "B-3".
Figures 6 and 7 show the test results of Vickers hardness tests of those. In the pictures
and the figures, "Front surface" indicates the surface in which electron beam wasdischarged and "Back surface" indicates the opposite side. In Figures 6 and 7, readdashed lines express the supposed center of EBW.
According to the results of the macrostructure tests and Vickers hardness tests,
it is found that heat-effected zone by EBW on the front surface is wider than that on theback surface because the front surface is a surface in which electron beam wasdischarged. Judging from the hardness gained from Vickers hardness tests, it is thoughtthat there is a correlation between "B" and the width of the non heat-affected zone "L",in which the hardness is almost the same as that of the base material. The values of "L"are described in Table 3. As shown in Table 3, the values of "L" become smaller as thoseof "B" become smaller.
Results of Charpy V-notch impact tests
Charpy V-notch impact tests were conducted with under-sized test specimens
whose thickness is 7.5mm. The temperature for the Charpy V-notch impact tests was
0°C, -30Υ and -60Υ. Figure 8 shows the results of the Charpy V-notch impact tests.
The significant difference in the test results at -60Υ and -30Υ among all test specimens
is not found and the values of the Charpy absorbed energy at -60Υ and -30Υ are verysmall as a whole. On the other hand, the difference in the Charpy absorbed energy at 0Υ
can be found depending on the type of the test specimens. The Charpy absorbed energyof the test specimens "B-13" whose "B" is 13mm is almost the same as that of the test
specimens "B-0" that is test specimens without EBW although the variation in test
results can be seen. The Charpy absorbed energy of the test specimens "B-4" in which
all over the V-notch area is a heat-affected zone is the smallest of all types of test
specimens. The Charpy absorbed energy of the test specimens "B-9" whose "B" is 9mm
exist between that of "B-4" and that of "B-0" or "B-13". The test results show that theCharpy absorbed energy depends on the width "B" and the wider "B" leads the higher
Charpy absorbed energy. Furthermore, the Charpy absorbed energy of the test
specimens "B-13" whose "B" is 13mm is almost same as that of test specimens without
EBW. Therefore, it is thought that the Charpy absorbed energy of test specimens madeby EBW may converge to that of test specimens without EBW when "B" is almost
13mm. This fact indicates the possibility that the Charpy absorbed energy can beevaluated adequately with the test specimens made with stop-hole-size cores by EBW.
Concluding remarks
This study examines the use of small steel pieces regarded as cores, which aretaken from stop-holes or bolt-holes. The targeted steel is the steel removed from a web
plate at a cross beam of steel girder bridges constructed in the 1920's or 1930's. Test
specimens for Charpy impact test are made with small steel pieces by Electron Beam
Weld (EBW) and the effect of the steel piece size on Charpy absorbed energy are䋭236䋭

examined.
As a result, it is found that the Charpy absorbed energy depends on the width of
the steel pieces and the wider steel pieces leads the higher Charpy absorbed energy and
it is thought that the Charpy absorbed energy of test specimens made by EBW may
converge to that of test specimens without EBW when the width of the steel pieces isalmost 13mm. This fact indicates the possibility that the Charpy absorbed energy can beevaluated adequately with the test specimens made with stop-hole-size cores by EBW.
Acknowledgments
This work was supported by the Grant-in-Aid for Scientific Research (S: No.
18106010) from the Japan Society for Promotion of Science (JSPS).
References
Japan Road Association (2007): Collection of Example of Retrofitting of Highway
Bridges. (in Japanese)
Japan Road Association (2009): Collection of Example of Retrofitting of Highway
Bridges. (in Japanese)
Chitoshi Miki and Takuyo Konishi (2007): Retrofit Engineering for Steel Bridge
Structures in Japan, A-0673, IABSE SYMPOSIUM, WEIMAR, 2007.
C. Miki, K. Ono, K. Yokoyama and T. Harada (2009): Identification of the causes offracture in the steel pipe column of a pedestrian bridge, Welding in the World,
Vol.53(9/10), pp.229-237.
Kenji Seo and Junichi Masaki (1982): Physical Interpretation for the Upper Shelf
Energy of Weld Zone in Charpy Impact Test, Journal of the Japan Welding
Society, Vol. 51, No.3, pp.39-45. (in Japanese)
Kenji Seo, Junichi Masaki, Fumio Nogata and Kunihiko Sato (1983): Effect of
Mechanical Heterogeneity on the Ductile-to-Brittle Transition Behavior of Weld
Metal in Charpy Impact Test, quarterly journal of the Japan Welding Society,Vol.1, No.2, pp.123-129. (in Japanese)䋭237䋭䋭242䋭

BRIDGES: THE NEXT GENERATION OF EXPERIMENTAL RESEARCHIN STRUCTURAL FIRESJohn L. Gross 1 and Stephen A. Cauffman 2
Abstract
Bridge fires, while not common, can have an enormous economic impact and canadversely affect the community served by the bridge, diminishing community resilience.
Regulatory approaches that have been successful for buildings may not be applicable tobridges. And, while provisions addressing the protection of the structural elements of abridge from high temperatures exist, little guidance is available to the bridge engineer toapply such provisions. This paper looks at these issues as well as a few examples ofrecent bridge fires, and attempts to make the case for the need to conduct large-scale
experiments on bridges and bridge components under realistic fire conditions. Such tests
can provide the technical basis for a performance-based approach to the design of bridges
to resist fires. A new structural fire test facility, under construction at the National
Institute of Standards and Technology, will enable the evaluation of the performance oflarge-scale bridge structures subjected to realistic fires. The capabilities of the National
Fire Research Laboratory are presented along with its technical specifications.
Introduction“Among the public safeguards that have been found necessary where buildings
are built in proximity to each other are those pertaining to fire. Such regulations arefounded on the long community experience that has been had with fires” [Ingberg, 1929].
Ingberg’s comment addresses, of course, the objectives of minimizing life and property
loss due to building fires. If, however, one were to extend his tenet to include
transportation systems, one would need to add the objectives of minimizing economic
impact and increasing community resilience. Highways are critical lifeline systems
enabling commerce and the transportation of people and goods within and amongcommunities. Failure of a single bridge can have a significant economic impact in termsof direct costs and lost productivity due to traffic disruptions and detours.
Regulations intended to reduce the risk of damage or collapse and minimize
economic impact by protecting bridge structures from high-temperature exposure do exist[NFPA, 2011], but guidance in their application is limited. A performance-based design
1

Associate Director of Structures Research, National Fire Research Laboratory, NIST, Gaithersburg, MD,
20899 USA2
Deputy Division Chief, Materials and Structures Division, NIST, Gaithersburg, MD 20899 USA䋭243䋭

approach for fire holds promise for bridge structures, but the technical basis forpredicting analytically the performance of a bridge structure exposed to fire, including
complex nonlinear behavior, is lacking. To date, there have been no fire tests on theperformance of bridge structures to provide the required technical basis and analyticalmodel validation.
Recent Bridge Fires
Garlock et al. [2011] have shown, through a literature review and case studies,that bridge fires can lead to significant structural damage, partial or complete collapse, as
well as costly detours in traffic flow. Garlock et al. go on to report on the frequency ofbridge fires and notes that a 2008 New York Department of Transportation national
survey found that, of the 1746 bridge failures reported (the majority resulting fromflooding or collision), 52 were due to fire and 19 due to earthquake; in other words,
bridges were almost three times more likely to collapse due to fire than to earthquakes.
In this paper two recent examples which led to significant cost of repair anddisruption are highlighted: the McArthur Maze fire which involved the collapse of asection of an overpass onto the roadway below, and the fire on the Mezcala (Mexico)cable-stayed bridge that involved the failure of one stay and damage to a second.
McArthur Maze Fire – Steel Plate Girder Failure
Early on the morning of April 29, 2007, a tanker truck carrying nearly 3400 L(9000 gallons) of gasoline overturned on the I-80/880 interchange in Oakland, California
and burst into flames [Bulwa and Firmrite, 2007]. The accident occurred at 3:41 A.M.
and just over 20 minutes later, at 4:02 A.M., the elevated roadway collapsed (see Figure1). “As the fire progressed, the bolted connections in the collapsed girder began toweaken due to heat and were placed under increasing load from the weakening of theremainder of the bridge” [Kodur, et al., 2010].

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Figure 1 - Collapse of the McArthur Overpass
Photograph by Robert Campbell (used with permission)
Mezcala Bridge Fire – Stay Cable Failure
In March of 2007, a traffic accident involving two school buses and a truck
transporting coconuts caused a fire on the Mezcala Bridge, a cable-stayed bridge in
Mexico, that resulted in failure of one stay and damage to a second [Zoli and Steinhouse,2007]. It is reported that flammability of the exterior corrosion protection system on thebridge cables added to the thermal loading on the cables.
Structural Fire Resistance Regulations – A Historical Perspective
A historical look at the development of regulatory approaches to fire resistance ofbuildings provides the context for discussion regarding the challenges faced indetermining the fire resistance of bridges.
Building Fires
In 1791, just four years after the U.S. Constitution was signed, President George
Washington issued the first regulations limiting building heights in the nation’s newcapitol of Washington D.C., “concerned as much about structural and fire safety as about
urban design” [Lewis, 1994]. Thus, regulations to limit the spread of fire dates back over
200 years.
The U.S. grew rapidly with population centers in Philadelphia, New York,
Boston, Chicago and San Francisco. Along with this growth came devastating fires that

䋭245䋭

affected all of these burgeoning cities. The Great Chicago Fire of 1871, in which over
17,000 buildings were destroyed, marked a significant turning point in building practicewhereby the fire resistance of a building began to be considered explicitly. “Although
the early knowledge of the requirements for fire protection resulted from a study of the
behavior of structures in fires and from examination of fire-damaged buildings, thedevelopment of skeleton-type construction made the necessity for fire endurance testing
apparent” [Shoub, 1961]. New York City was the first American city to introduce a
standard fire test method in the 1899 New York Building Code [Babrauskas andWilliamson, 1978]. But it was the Great Baltimore Fire of 1904 that prompted theAmerican Society for Testing and Materials (ASTM) to organize a national effort tostandardize fire resistance testing; “A Standard Test for Fire-Proof Floor Construction”
was issued just three years later in 1907. ASTM continued its work and in 1917
Committee C-5 issued a report that proposed a standard test method notable for two
aspects, “the provision that structures be classified by their attained fire resistance”[Shoub, 1961], and the notion that “a furnace does not heat up instantaneously; forreproducible results, this initial heating should be quantified” [ Babrauskas andWilliamson, 1978]. The furnace time-temperature curve introduced by Committee C-5
remains unchanged to this day in the ASTM Standard E 119, Standard Test Methods forFire Tests of Building Construction and Materials [ASTM, 2011], the standard
commonly used in the United States for establishing fire resistance ratings.
It was Simon Ingberg of the National Bureau of Standards, now the National
Institute of Standards and Technology( NIST), who introduced the fire “severity” concept
suggesting that “all fires of the same severity have approximately the same effect on a
structure” [Babrauskas and Williamson, 1978]. “Ingberg’s work on fire severity and fire
resistance was adopted by national standards and model building codes…and hisclassification of building types [fireproof, incombustible, exterior-protected and wood]remains the basis for requirements for fire resistance of building components…” [Evanset al., 2001]. Fire resistance ratings for buildings, as established by ASTM E 119, areintended to ensure the that a fire will not spread beyond the compartment of origin.
One significant addition to building codes in the United States has been theprovision that “an approved automatic sprinkler system…shall be allowed to be
substituted for 1-hour fire-resistance-rated construction” [ICC, 2006]. This provision,
known as the “sprinkler tradeoff” provision, recognizes the effectiveness of automatic
sprinklers in suppressing a fire in its early stages, thereby preventing the fire frombecoming a threat to the building or its occupants.

䋭246䋭

Bridge Fires
Consideration for the need for fire protection for bridges is more recent. And, as
noted in the introduction, the impact of a bridge fire extends beyond life safety andreplacement cost to the broader impact on the community served by the bridge.
The National Fire Protection Association (NFPA) Standard for Road Tunnels,
Bridges, and other Limited Access Highways [NFPA, 2011] contains provisions for theconsideration of fire. Quoting directly from Chapter 6:
6.3.1 …all primary structural elements shall be protected in accordancewith this standard in order to:(1) Maintain life safety(2) Mitigate structural damage and prevent progressive structural
collapse(3) Minimize economic impact
6.3.2 Critical structural members shall be protected from collision andhigh temperature exposure that can result in dangerous weakening orcomplete collapse of the bridge or elevated highway.
However, as pointed out by Garlock et al. [2011], “…current bridge design codes andstandards offer limited information concerning the fire hazard.”
A Case for Structural Fire Testing of Bridges
From the above discussion, one can identify three regulatory approaches to fire
resistance for buildings: (1) construction restrictions (e.g., zoning and occupancy
restrictions), (2) limiting the spread of fire beyond the compartment of origin (e.g., fire
resistance rating requirements), and (3) active fire suppression (e.g., sprinklers). It is not
at all clear, however, that any of these regulatory strategies would be effective in
reducing the hazard from fire for bridges and elevated highway structures.
Additionally, results of tests on building components and assemblies are generally
not applicable to bridge structures. For example, building fires are generally considered
to be confined to a compartment, while bridges are likely to be exposed to plume fires.
The combustible materials in a building (e.g., wood, paper, upholstery) are likely to havedifferent burning characteristics than those affecting a bridge which may involve
petroleum products (hydrocarbons). Finally, structural elements of a building are often
different from elements found in bridges (e.g., steel plate girders or box sections, prestressed concrete beams, suspension cables, etc.). No systematic study has been

䋭247䋭

conducted to date to quantify the effects of damaging hydrocarbon pool fires on the wide
variety of bridge forms in use today. Such a study is required to fully understand the
behavior and modes of failure exhibited by the various structural forms when exposed to
extreme temperatures.
By treating fire as a design condition in the design and analysis of bridges, one
can implement a performance-based design approach. Performance-based design
generally involves the calculation of the structural response of a bridge to the effects of a
postulated fire, developed through consideration of possible fire scenarios. One of themain obstacles for moving towards performance-based fire safety design is the lack ofknowledge about bridge fire response. Numerical models are needed as well as design
tools, but these models and tools need to be validated with experiments since the bridge
response under high thermal loads is complex and nonlinear.
To conduct such experiments requires a facility with the capability to:x Conduct tests on real-scale structural systems and components
x Apply controlled loads to test structures to simulate true service conditions
x Create realistic fires that grow, spread and decay
x Characterize the fires in real timex Measure the response of structural systems and components to the point of
incipient collapse.
Until now, no laboratory in the world has possessed this combination of capabilities.
The National Fire Research Laboratory (NFRL)The National Institute of Standards and Technology (NIST) is adding a new,unique facility that will serve as a center of excellence for fire performance of structures
ranging in size from small components to large systems (see Figure 2). The laboratory,
called the National Fire Research Laboratory (NFRL), will be led, managed and operated
as a collaborative facility through a public-private partnership between NIST andindustry, academia, and other government agencies.
The work of the laboratory will be focused on the NIST Engineering Laboratory
mission: to promote US innovation and industrial competitiveness in areas of national
priority by anticipating and meeting the measurement science and standards needs fortechnology-intensive manufacturing and construction in ways that enhance economic
prosperity and improve the quality of life.

䋭248䋭

Figure 2 - Rendering of the new NIST facility for large-scale structural fire research
Scientists and engineers from industry, academia, and government agencies will
work side-by-side with NIST researchers to address significant technical problems andfill critical knowledge gaps, and international scientists and engineers will be welcome topartner with NIST in areas of mutual interest. Projects may be funded by industry andgovernment on a cost-shared basis.
The additional capabilities will allow NIST to:
x

x

x
x

Test the performance of large-scale structures, including bridge components,
subjected to realistic fires and structural loading under controlled laboratory
conditions.
Develop an experimental database on the performance of large-scale structuralconnections, components, subassemblies, and systems under realistic fire andloading.
Validate physics-based models to predict fire resistance performance ofstructures.
Enable performance-based standards for fire resistance design of structures andfoster innovations in design and construction.

The NFRL is adding 1990 m2 (21,400 sq ft) laboratory space to its existing Large Fire
Laboratory (Building 205) (see Figure 3) and installing an environmental control system(ECS) to supplement the existing ECS to accommodate fires up to 20 MW heat release
rate.

䋭249䋭

Figure 3 -Floor Plan of the National Fire Research Laboratory Expansion
The new laboratory space will accommodate structural systems or components
9 m (30 ft) high and roughly 12 m (40 ft) by 18 m (60 ft) in plan. Gravity loading will beapplied using hydraulic actuators or fixed loads. Fully involved building fires, fueled bygas or liquid fuel, wood cribs, or actual building contents, will be employed to simulatebuilding fire conditions. Characteristics of the fire (heat release rate) will be measured
accurately using calorimetry.
The test area will consist of a 18.3 m u 27.4 m (60 ft u 90 ft) strong floor withanchor points on a 0.61 m u 0.61 m (2 ft u 2 ft) grid. The floor will be supported on anine-cell reinforced concrete box girder providing a basement below the strong floor witha ceiling height of 2.7 m (9 ft). To one side of the strong floor will be a 9.1 m (30 ft) highu 18.3 m (60 ft) wide concrete strong wall with anchor points on the same grid as thestrong floor. The strong wall will act to stabilize a test specimen to prevent uncontrolled
failure, provide lateral restraint, or to laterally load a structure to simulate earthquake
damage. A 13.7 m u 15.2 m (45 ft u 50 ft) hood, centered above the strong floor, willcapture and remove smoke and hot gases.
The size of the test area was selected to enable the testing of large-scale structural
systems or components; including bridge girders, cable systems, piers, etc.; underrealistic hydrocarbon fires and controlled loading comprised of self-weight and vehicles.
Technical specifications for the NFRL expansion are given in Appendix A.

䋭250䋭

Summary
In summary, the following points have been made:x Fire represents a significant hazard in bridges
x Consequences of a bridge fire include:o structural damage, partial or complete collapse necessitating repair or
replacemento disruption to traffic flow during repair or replacemento economic loss and adverse impact on community resiliencex The 100-plus year history of fire resistance regulations for building structuresoffers little guidance for bridge and transportation structuresx Current bridge design codes and standards offer limited information concerning
the fire hazard
x Fire hazard can be overcome by addressing fire explicitly in the design andanalysis of bridges
x Identification of failure modes and validation of advanced numerical models
requires well controlled full-scale experiments
x The National Institute of Standards and Technology’s new, unique fire/structure
test facility, the National Fire Research Laboratory, will enable the evaluation of
the performance of large-scale bridge components subjected to realistic fires.
References
ASTM (2011). Standard Test Methods for Fire Tests of Building Construction andMaterials, ASTM E 119 11 a, ASTM International, Conshohocken, Pa.
Babrauskas, V. and Williamson, R. B. (1978). “The Historical Basis of Fire Resistance
Testing – Part II,” Fire Technology, Vol. 14, No. 4, Nov.
Bulwa, D. and Firmrite, P. (2007). “Tanker Fire Destroys Part of McArthur Maze,” San
Francisco Chronicle, California.
Evans, D., Gross, D. and Wright, R. (2001). “Test of the Severity of Building Fires,” A
Century of Excellence in Measurements, Standards, and Technology, Lide, D.R.
(ed.), NIST Special Publication 958, National Institute of Standards andTechnology, Gaithersburg, MD.
Garlock, M., Payá-Zaforteza, I., Kodur, V., Gu, L. (2011). “Fire Hazard in Bridges:
Review, Assessment and Repair Strategies” (submitted for publication in
Engineering Structures, Elsevier).

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ICC (2006). International Building Code, International Code Council, Falls Church, VA.
Ingberg, S.H. (1929). “Fire Resistance Requirements in Building Codes,” Quarterly of
the National Fire Protection Association, Vol. 23, No. 2, October 1929, NFPA,
Boston, MA.
Kodur VR, Gu L, Garlock M.E. (2010). “Review and Assessment of Fire Hazard in
Bridges” Journal of the Transportation Research Board No. 2172, Transportation
Research Board of the National Academies, Washington D.C. (USA)
Lewis, R. (1994). “Testing the Upper Limits of D.C. Building Height Act,” The
Washington Post, Saturday, April 23, Washington, D.C.
NFPA (2011). Standard for Road Tunnels, Bridges, and other Limited Access Highways,
AFPA 502, 2011 edition, National Fire Protection Aassociation, Quincy, MA.
Shoub, H. (1961). “Early History of Fire Endurance Testing in the United States,”Special Technical Publication No. 301, ASTM, Conshohocken, PA
Zoli, T. P., and Steinhouse, J. (2007). “Some Considerations in the Design of Long Span
Bridges against Progressive Collapse,” Proceedings of the 23rd U.S.-Japan Bridge
Engineering Workshop, Tsukuba, Japan, Nov. 5-7
Appendix A - Technical Specifications

Strong Floor
x 18.3 m u 27.4 m (60 ft u 90 ft) post-tensioned floor with full basementx 9 cell RC box girder with 406 mm (16 in) thick shear walls at 3.0 m (10 ft) o.c.
x Basement ceiling height: 2.7 m (9 ft)

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x
x
x
x
x

Floor thickness: 1.07 m (3 ft-6 in) with 152 mm (6 in) sacrificial top surface
1218 anchor points on 0.61 m u 0.61 m (2 ft u 2 ft) grid (sleeves or anchors)
Load per anchor point: 445 kN (100 kip) up or down
Shear capacity per anchor point: 222 kN (50 kip) (at top of slab)
Moment capacity per anchor point: 136 kN·m (100 ft kip) (at c.g. of strong floor)

Strong Wallx 9.1 m high u 18.3 m wide (30 ft high u 60 ft wide)
x 1.2 m (4 ft) deep post-tensioned concrete wall
x 420 anchor points on 0.61 u 0.61 m (2 ft u 2 ft) gridx Horizontal load: 146 kN/m (10 kip/ft) at 9.14 m (30 ft)
ECS Hood and Pollution Control System
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x Height of rails above floor: 11.2 m (36 ft-8 in)x Clearance, bottom of bridge-to-floor: 9.8 m (32 ft)
Configurable Hydraulic Loading System
x Hydraulic Power Unit 340 L/min (90 gal/min)
x Actuators (double acting) 762 mm (30 in) stroke w/ servo valve, load cell and swivels
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x Four hydraulic service manifolds
x Controller

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Test construction for confirmation of application to the new blast method using
the special tool on steel bridge coating repair construction site
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FHWA’S NEW INFRASTRUCTURE RESEARCH AND TECHNOLOGYSTRATEGIC PLAN
Ian M. Friedland, P.E. 1

Abstract
In the fall of 2010, the Federal Highway Administration (FHWA) undertook aneffort to develop a new, longer-term strategic plan and roadmap for infrastructure
research and development, and technology deployment. This paper provides an overview
of the new infrastructure R&D strategic plan, and deployment roadmap.
Background
Over the years, the Federal Highway Administration’s (FHWA) bridge andstructures research and development (R&D) efforts have addressed a range of topics andfocal areas. Past efforts have included research as wide-ranging as that of assessing andreducing the seismic vulnerability of the U.S. national highway system, to curved steel Igirder bridge behavior, to understanding the forces associated with spill-throughabutment scour, to the use of ground penetrating radar for concrete bridge deck
inspection. R&D topics were typically developed based on the knowledge of technology
gaps by FHWA engineers, input from a wide variety of stakeholders, and–occasionally–
direction from the U.S. Congress. In addition, “failures” within the highway
infrastructure also drove the decision-making regarding bridge and structures R&D.
Starting in the middle of 2006, the FHWA Office of Infrastructure R&D took amore strategic approach to its funded and planned research agenda. This approach
looked at blending the two distinct infrastructure program areas – pavement R&D, andbridge and structures R&D – into a more comprehensive and integrated infrastructure
R&D program. The overall focus of this effort was on long-term infrastructure
performance. The result of this was a multi-tiered, multi-year strategic plan and researchroadmap, as documented in the 2008 FHWA publication titled “Highways of the Future:
A Strategic Plan for Highway Infrastructure Research and Development” (publication
FHWA-HRT-08-068, July 2008).
The 2008 strategic plan addressed six primary R&D areas:xx•Long Term Infrastructure Performance
Durable Infrastructure Systems
Accelerated Highway Construction

1

Assistant Director for Bridge and Structures R&D, Federal Highway Administration,
Washington, DC, USA

䋭267䋭

x
x
x

Environmentally Sensitive Infrastructure
Performance-Based Specifications
Comprehensive & Integrated Asset Management

The basis for focusing on these six areas are the many critical challenges currently facing
highway agencies, not only in the United States, but worldwide, including:xxx

The need to extend the service life of existing highway infrastructure.
The need to build, rehabilitate, and rebuild infrastructure in ways that:o Minimizes the impact of construction activities on already congested highways.
o Optimizes the overall cost/benefit for the improved infrastructure.
o Facilitates future adaptation to accommodate changing demands.
The need to effectively address the mobility challenges posed by natural or man-madeextreme events and hazards – including earthquakes, hurricanes, floods, collisions,and acts of terrorism – by designing and constructing less vulnerable infrastructure tominimize loss, and employing rapid restoration techniques to restore functionality
after a disaster occurs.

This FHWA infrastructure R&D strategic plan was the guiding document for all
infrastructure R&D investments between 2007 and 2010.
The 2011 FHWA Infrastructure Research and Technology Strategic Plan
In the fall of 2010, FHWA initiated an effort to better integrate all elements of theagency’s work related to highway infrastructure; i.e., not just R&D, but inclusive of all
highway infrastructure technology and innovation development and deployment
including design and construction, structures and pavement, asset management and longterm infrastructure performance. This required a more inclusive outreach within theagency which brought together all of FHWA’s infrastructure offices and functions. The
result was a new multi-year strategic plan and roadmap that describes the direction andoutcomes to be pursued through FHWA’s infrastructure Research and Technology (R&T)
program for the next five to ten years. It is founded on and informed by input from a
broad array of highway stakeholders gathered through both formal and informal
mechanisms.
The United States highway system is entering an unprecedented era of change thatbrings with it significant challenges and opportunities for highway infrastructure. The
country is experiencing unprecedented fiscal challenges to operate, maintain, and invest
in infrastructure in order to maintain a state of good repair and provide the improvements
needed to sustain and grow the economy. In these challenging times, there is no choice
but to develop and apply innovative solutions to meet the needs of the travelling public.

䋭268䋭

Past infrastructure research and technology investments have produced and put
into practice innovations that have resulted in longer-lived assets at lower costs, reduced
environmental impacts, saved lives, and improved economic efficiency. Additionalinnovation will be needed to further improve safety, reduce congestion, addressenvironmental and energy concerns, provide the quality highway system the nation’s
citizens expect and address the Department’s Strategic Goals. Conducting research that
addresses national highway infrastructure needs, developing new and updated policy,
guidance and technologies to address these needs, and effectively deploying policy,
guidance, and technologies are key programmatic activities that must be embraced tomeet the challenges faced by the United States – and similarly worldwide. These efforts
will be directed to ensure that highway infrastructure is delivered in ways that impact
traveler mobility as infrequently as possible, for the shortest amount of time, providing
the greatest mobility and safety. Transportation infrastructure must be managed in a waythat addresses the new challenges and puts the needs of the American people and theircommunities first.
This overarching FHWA R&T infrastructure strategic plan and its supporting
roadmap is now guiding FHWA’s Infrastructure R&T efforts for the next 5 to 10 years.
The plan provides a comprehensive focus and direction across organizational boundaries;
assists in prioritizing program initiatives, allocating resources, and improves theprocesses relative to how FHWA achieves its mission well into the future; and recognizes
the inter-relationships and interdependencies among the different infrastructure
disciplines and provides a framework for collaboration across disciplines and with other
FHWA programs.
The strategic direction articulated in this plan is founded in the FHWA Strategic
Plan and the U.S. DOT Strategic Plan, FY 2010- FY 2015, Transportation for a New
Generation 2. The outcomes that will be achieved to support the goals identified in theseplans are as follow:
1

1.
2.
3.
4.
5.
6.
7.

Highway safety is improved;
Management of the infrastructure system is continuously improved;
Economic returns on transportation infrastructure investments are improved;
Delivery of high quality infrastructure projects is expedited;
Durability and longevity of highway infrastructure are improved;
The condition of the highway infrastructure is improved;
The sustainability of highway infrastructure design, construction, maintenance andoperation is improved and adverse environmental impacts are reduced throughenvironmental stewardship; and8. Personal and commercial mobility is improved.
The specific objectives that will be pursued to achieve these outcomes andcontribute to achievement of the FHWA and DOT strategic goals are as follows:

䋭269䋭

1. Reduce the number of fatalities attributable to infrastructure design characteristics
and work zones.
2. Improve the safety and security of highway infrastructure.
3. Improve the management of infrastructure assets and advance the implementation ofa performance-based program for the NHS.
4. Improve the ability of transportation agencies to deliver projects that meet
expectations for timeliness, quality and cost.
5. Reduce user delay attributable to infrastructure system performance, maintenance,
rehabilitation and construction.
6. Improve highway condition and performance through increased use of design,materials, construction and maintenance innovations.
7. Reduce the life-cycle environmental impacts of highway infrastructure (design,
construction, operation, preservation, and maintenance).
The outcomes delivered through the objectives articulated in this strategic plan
are aimed at delivering benefits to the American public by enabling improvements insafety, performance, and cost effectiveness of the U.S. highway infrastructure, while
minimizing the environmental impacts of highway construction, maintenance, andrehabilitation. The results will make possible reductions in highway congestion,
improved travel time reliability, improvements in highway safety, and enhancement of
the overall driving experience for the American public.
FHWA Infrastructure R&T Program Roadmap – Objectives, Strategies, andInitiatives
Details on the FHWA Infrastructure R&T Strategic Plan are provided in thefollowing program roadmap. As noted earlier, this plan and roadmap will drive FHWA’s
infrastructure R&D investments for the next 5 to 10 years.
Objective 1: Reduce the number of fatalities attributable to infrastructure design
characteristics and work zones.
Strategy
1.1 Develop and deploy bestpractices and opportunities
to improve infrastructure
safety performance.
1.2 Develop and deploy
technologies, standards, andtest methods that optimize
surface characteristics withregard to friction, texture
and splash and spray.

Initiative
1.1.1 Assessment of the adequacy of current policies andpractices with regard to ensuring the safety of highway
infrastructure and provide updated policy and bestpractice where warranted.
1.2.1 Develop, standardize, and deploy test methods,
technologies and specifications for enhancing
friction/texture on new and existing surfaces.
1.2.2 Identify and prioritize data needed to fully
characterize the safety-related characteristics of
highway infrastructure, including geometrics and surface

䋭270䋭

1.3 Develop and deploy
technical guidance tosupport infrastructure safety
management programs.

1.4. Develop and deploy
construction administration
practices that enhance safe
operation of the highway
system by reducing workzone exposure.

characteristics.
1.2.3 Evaluate, refine, and standardize test methods forfriction and texture for effectiveness, repeatability, andreproducibility.
1.3.1 Improve effectiveness of friction management
programs.

1.4.1 Evaluate existing practices for accelerated
construction practices related to contract administrationfor effectiveness to reduce crash risk in work zones.
1.4.2 Update and enhance policy and technical guidancewith regard to accelerated construction practices related
to contract administration.
1.4.3 Develop and deliver training and technical support
for implementation of policy and technical guidance
related to accelerated construction and contract
administration.

Objective 2. Improve the safety) and security of highway infrastructure.
Strategy
2.1 Develop and deploy
planning, analysis and design
methodologies for highway
infrastructure to reduce
vulnerability to physicaldamage. (Design)
2.2 Develop and deploy
hazard mitigation,
adaptation and restoration
strategies and techniques.
(Rehab)2.3 Develop and deploy
improved decision support
tools and sensing andmonitoring technologies forhazard detection.

Initiative
2.1.1 Implement an integrated interagency Federal
approach that consolidates capabilities in a unified effort
for security to prevent human induced hazard events.
2.1.2 Improve and optimize Infrastructure System design
to improve safety and security.
2.2.1 Research, develop, and deploy strategies,
technologies, and programs to mitigate specific known
infrastructure hazards and adaptation needs.
2.2.2 Development of hazard mitigation and adaptation
countermeasures for existing and new structures that
includes a national data archive resource that captures
experimental data as well as field reconnaissance data.
2.3.1 Development of improved decision support tools
and methodologies for assessing hazards toinfrastructure.
2.3.2 Research, develop, and deploy better detection
and surveillance technologies for evaluation, prevention,

䋭271䋭

(Monitoring andassessment)

and mitigation of extreme events.

2.4.1 Development of protocols/inspection techniques
for rapid assessment of infra structures after a hazard
2.4 Develop and deploy
event.
methodologies and guidance
2.4.2 Research and deployment of ‘smart structure’for assessing safety ofdesigns that sense damage and provide active/semiinfrastructure after a hazard
active control of structural response to hazard events.
event. (Repair and/or
2.4.3 Research, develop, and deploy new infrastructure
replace)systems that can rapidly be repaired and reconstructedfollowing a hazard event.
Objective 3: Improve the management of infrastructure assets and advance theimplementation of a performance-based program for the NHS.
3.1.1 Establish performance standards forinfrastructure
3.1.2 Identify and fill gaps in performance prediction
models and practices across the asset life-cycle
3.1 Develop and deploy
spectrum.
reliable performance
3.1.3 Develop and deploy improved traffic prediction
prediction models andtools
practices in the design,
3.1.4 Develop and deploy improved prediction tools forconstruction andenvironmental and climate change impacts on
management of the highway
infrastructure performance.
infrastructure.
3.1.5 Conduct research to understand performanceissues; and identify design, construction, preservation,and maintenance strategies to assure durability andlongevity of infrastructure.
3.2.1 Develop improved methodologies & tools to more
effectively collect and manage infrastructure condition
3.2 Develop and deploy sound data at the national, network and project levels tomore effectively identify causes of deterioration and tomeasures and practices tomake more informed decisions on system performanceassess infrastructureand health.
condition and to assure data
quality in infrastructure
3.2.2 Develop and deploy non-destructive evaluation
management and(NDE) tools, software, and guidance for condition
performance predictions.
evaluation, construction quality assurance, and theprediction of key highway infrastructure performanceat the network & project levels through the entire life

䋭272䋭

cycles of the infrastructure assets.
3.2.3 Conduct research to develop and deploy new
methods & technologies to monitor infrastructure
conditions in real time.
3.2.4 Develop tools and technologies for, and promoteconsideration of uncertainties and risks in, data quality.
3.2.5 Develop and implement guidance, technicalsupport, training, and technology transfer for data
quality management and to control affects from data
variability.
3.3.1 Identify, assemble, analyze the required dataneeds & develop analysis methodologies for an overall
integrated asset management approach.
3.3 Develop and deploy
3.3.2 Develop and deploy tools, technologies, anddecision support tools,
systems for project and program level management
systems, and processes todecisions
support rational, andcomprehensive engineering
3.3.3 Develop and deploy tools and technologies forand economic analysis
cross-asset analysis involving existing assets andmethods for project,
proposed future assets, and including the inherent
program, and national level
uncertainties and risks.
investment decisions.
3.3.4 Develop tools and technologies for, and promoteconsideration of, uncertainties and risks inherent inlonger term investment decision making process.
3.4.1 Define & deploy a consistent and reliable method
3.4 Develop and deploy
to describe and project the health of the highway
guidance, management
system in collaboration with stakeholders.
approaches, and policies for3.4.2 Conduct research to determine the impact
management of infrastructure
infrastructure condition has on users, and develop
assets and foranalysis tools to better assess those impacts.
implementation of a
3.4.3 Develop and deploy management approaches andperformance-based program
policies for corridors, tunnels, ancillary structures andfor infrastructure on the NHS.
other transportation assets.
Objective 4: Improve the ability of transportation agencies to deliver projects that meet
expectations for scope, timeliness, quality and cost.
Strategy
4.1 Develop and deploy
expanded and consistent useof the elements of a quality
assurance program to

Initiative
4.1.1 Develop and deploy a composite index for projectvalue that incorporates scope, costs, time, and quality.
4.1.2 Develop guidance and tools, conduct program
reviews, and deploy effective Quality Assurance plans

䋭273䋭improve infrastructure
design, materials testing,construction, and inspection
procedures.
4.2 Develop and deploy
innovative processes andproject management
practices to enhance projectdelivery in highway designand construction.
4.3 Develop and deploy
contracting tools andpractices to effectively
manage risk in acceptance ofand payment for construction
and materials.

and best practices.

4.2.1 Identify, further develop, and deploy improved,
streamlined and refined analysis, design, andconstruction procedures.
4.2.2 Develop and deploy tools and guidance for innovative
project delivery methods.

4.2.3 Develop and deploy tools and technologies toassure and improve the quality & maintainability ofaccelerated construction projects.
4.3.1 Develop and deploy tools that project the long-term
performance and value as a result of the as-built condition of
the infrastructure asset.

4.3.2 Develop and deploy approaches and specifications
that link payment and acceptance to long-term
performance and quantities to minimize agency risk.

Objective 5. Reduce user delay attributable to infrastructure system performance,
maintenance, rehabilitation and construction.
Strategy
5.1 Develop and deploy tools
and methodologies to assess
the impact of decisions(design, construction,
contracting, etc.) on user
delay.

5.2 Develop and deploy
construction, inspection,
maintenance, preservation,and rehabilitation practices
that minimize impact tousers.

Initiative
5.1.1 Improve tools to more effectively assess user
delay during design.
5.1.2 Develop and deploy tools to evaluate the impacts
of different construction delivery approaches on user
delay (i.e. full closure vs. variable closures) at the
project and corridor levels.
5.1.3 Develop methods to evaluate the impacts ofdifferent contracting methods on user delay (i.e. designbuild, warranties).
5.2.1 Develop and deploy techniques for prefabricatedconstruction of the infrastructure.
5.2.2 Research, develop and deploy technologies andprocesses to accelerate construction and preservation.
5.2.3 Develop and deploy approaches to automate theconstruction and preservation inspection, sampling andtesting associated with the production and placement of
highway related materials and systems (includes all
related highway systems; pavement, bridge, culvert,

䋭274䋭

etc.).
5.2.4 Research, develop and deploy new approaches to
conduct infrastructure condition inspections tominimize user delay.
Objective 6: Improve highway condition and performance through increased use ofdesign, materials, construction and maintenance innovations.
Strategy
6.1 Develop and deploy
approaches to effectively andsystematically preserve andimprove the highway
infrastructure condition andperformance.

6.2 Develop and deploy
design and preconstruction
technologies and innovations
to improve infrastructure
condition, durability andservice life, andconstructability.
6.3 Develop and deploy
alternative project delivery
methods, construction
approaches andspecifications where theemphasis is on the long-term
performance of theinfrastructure system.
6.4 Develop and deploy
methods that will improve
the quality of materials andsystems used for highway
infrastructure.

Initiative
6.1.1 Develop methodologies to effectively apply
treatments using asset management principles at the
project and network levels.
6.1.2 Develop and deploy technologies and processes toenhance the effectiveness of preservation activities.
6.1.3 Develop and deploy technologies and techniques
to enhance the effectiveness of rehabilitation andreconstruction activities.
6.2.1 Deploy existing and proven criteria, technologies
and procedures to design infrastructure to more reliably
achieve intended performance and service life.
6.2.2 Research and develop the next generation ofanalysis and design tools to improve design reliability toachieve intended performance and service life.
6.2.3 Develop and deploy contracting procedures toimprove infrastructure condition and performance.
(warranties, value engineering, etc. to allocate risk)
6.3.1 Develop and deploy tools to assist in the
allocation of risk between agencies and industry.

6.3.2 Develop and deploy technologies to improve theefficiency of construction processes.

6.4.1 Develop and deploy new tests and methods tomore accurately and efficiently characterize materials.
6.4.2 Develop and deploy best practices for selection ofmaterials and systems, tests and methods to assure andimprove quality.
6.4.3 Develop and deploy innovative materials, andsystems to improve the durability and longevity of䋭275䋭highway infrastructure.

Objective 7: Reduce the life-cycle environmental impacts of highway infrastructure(design, construction, operations, preservation, and maintenance).
Strategy

7.1 Advance the applicationof sustainable practices inproject level infrastructure
design.

7.2 Develop and deploy
sustainable methods toreduce air pollutants andother emissions resulting
from the construction andpreservation practices.
7.3 Develop and deploy
sustainable methods toreduce water runoff andpollutants throughimprovements in design,
construction, operations,maintenance, andpreservation.
7.4 Develop and advancesustainable practices that
reduce noise during

Initiative
7.1.1 Develop and deploy Life-Cycle Assessment (LCA)
methodologies, tools, policy and guidance to implement
and to quantify the environmental impacts of design.
7.1.2 Develop and deploy tools and practices forassessing sustainability in project-level decision-making.
7.1.3 Develop policy and guidance, and implement
sustainable design approaches to adapt to the impactsof Climate Change.
7.1.4 Develop policy and guidance, and implement
context sensitive design practices that improvesustainability.
7.1.5 Evaluate, develop policy and guidance, andimplement sustainable aspects of existing and newmaterials, technologies and practices.
7.1.6 Develop policy and guidance, and evaluate
practices that maximize the use of marginal and locallyavailable materials.
7.2.1 Evaluate, develop policy and guidance, andimplement aspects of infrastructure materials andtechnologies to reduce air pollutants and otheremissions.

7.3.1 Evaluate, develop policy and guidance, andimplement use of materials and technologies to reducewater runoff and pollutants.
7.3.2 Develop policy and guidance, and implement
practices to improve storm water management.
7.3.3 Develop policy and guidance, and implement
practices related to non-storm water runoff andpollutants.
7.4.1 Develop, standardize and deploy test methods forpavement noise to more effectively characterize noise
performance of surfaces.

䋭276䋭construction and throughout
the service life.

7.4.2 Evaluate, develop policy and guidance, andimplement use of materials and technologies that
reduce noise such as optimizing pavement surface mix
designs to reduce noise.
7.5.1 Evaluate, develop policy and guidance, and
7.5 Identify, develop, andimplement aspects of materials and technologies for theadvance alternative energy
use of alternative energy sources.
sources for use during
construction, operations,7.5.2 Develop policy and guidance, and implement
preservation andinfrastructure component technologies that generate
maintenance.
energy.
7.6 Advance and increase the 7.6.1 Evaluate the long-term performance of renewable,
use of renewable, reusable,
reusable, and recycled (3R) materials and implement theand recycled (3R) materials in expanded use of 3R materials in highway-related
highway-related
infrastructure.
infrastructure.
7.7.1 Develop, evaluate, improve, and deploy practices
7.7 Minimize impacts ofthat minimize impacts on wildlife.
highway infrastructure on
7.7.2 Develop and deploy highway materials that are
wildlife.
benign or less damaging to wildlife.

References
1

FHWA Strategic Plan. Washington, DC : U.S. DOT, Federal Highway Administration,
October 2008. Publication No. FHWA-PL-08-027.

2

U.S. DOT Strategic Plan, FY 2010- FY 2015, Transportation for a New Generation,
Washington, DC : U.S. Department of Transportation, April 15, 2010.

䋭277䋭

䋭278䋭

Page Left Blank

䋭280䋭

UNBONDED PRESTRESSED COLUMNS FOR EARTHQUAKERESISTANCE
Alex S. Larkin1, David H. Sanders2, M. Saiid Saiidi3
Abstract
The implementation of unbonded post-tensioning in bridge columns reduces
damage and repair time by minimizing residual displacements. This research
investigates the use of unbonded post-tensioned tendons in bridge columns. Twounbonded post-tensioned columns were tested that are identical except for the amountof longitudinal reinforcement crossing the joint between the footing and column base.
The placement of post-tensioning in a full-scale column was taken into consideration
which resulted in four tendons being placed around the center of the column crosssection instead of one through the center of the column as has been done in previous
experiments. The tendons are anchored in the side of the footing for ease of
replacement following an earthquake.
Introduction
The number one concern for engineering seismic design is to maintain life
safety. Once a structure can maintain life safety through an earthquake, the next stepis to minimize the amount of damage to the affected structure. Minimizing the amountof damage will allow for rapid repairs and minimal closure time. For bridge columns,
the use of unbonded post-tensioned tendons reduces the amount of residual
displacement following an earthquake, allowing for minimal closure time of thebridge while repairs are made.
For full-scale columns, the amount of post-tensioning needed to promote recentering effects would be between 8% and 10% f’cAg. Typically, the initial stress inthe tendons is between 20% to 30% fpu. Therefore, if the column is 60 inches (1524mm) in diameter, and 0.6 in (15.2 mm) strands are used, a total of between 62 and 86
strands would be needed. In addition to this being a large number for one tendon,
multiple tendons allows for a tendon to be replaced while still maintaining posttensioning and also increases redundancy. The columns being tested are 0.4-scalemodel, with a diameter of 24 in (610 mm). Therefore, to evenly distribute the force
required for re-centering, four tendons, each with four 0.6 in (15.2 mm) Grade 270 ksi______________________
1
Graduate Research Assistant, University of Nevada Reno
2
Professor, University of Nevada Reno
3
Professor, University of Nevada Reno

䋭281䋭

(1862 MPa) 7-wire strands were evenly spaced at 5.4 in. (137.2 mm) around thecenter of the column cross section (see Fig. 1).
Following a major earthquake, bridge columns are likely to have undergone
large lateral displacements. Post-tensioned columns are typically designed to not havetendons yield or be damaged. It is critical for it to be possible to inspect and replace
the tendons. Previous tests have utilized straight tendon or bar that exited through thebottom of the footing; this configuration makes it impossible to replace the tendon. By
exiting out the side, or out the top of the footing for large footings (180 degree bend
for the post-tensioning tendon), the tendons can be removed. For these specimens, thetendons exited out the side of the footing (see Fig. 2).
Literature Review
Excessive residual displacement is an issue with columns following anearthquake. Unbonded post-tensioned tendons have been used in columns that haveshown reductions in residual displacements. Research performed by Ou showed there-centering capabilities of unbonded post-tensioned columns. Ou’s report also
indicates that high lateral column displacements lead to high strain in the unbonded
post-tensioned tendons (Ou et. al. 2009). Hewes has also indicated the importance inthe selection of the initial post-tensioned force in the tendon. The higher tendon forces
aid in reducing the residual displacements at low drift levels, but can lose theireffectiveness at high drift ratios due to the yielding of the tendon. Hewes’ report also
indicates the benefit of using unbonded tendons as opposed to bonded tendons
because the localized inelastic straining due to large drifts can be avoided,maintaining the re-centering force throughout testing or throughout an actual
earthquake (Hewes and Priestley 2002). While the column may still show damage
depending on the intensity of the earthquake, repairs can be made quicker when thecolumn re-centers itself. Sakai concluded that adding a steel jacket to the column
plastic hinge region as well as locally unbonding the mild reinforcing steel that crossthe joint between the footing and column base prevented damage compared to aconventionally reinforced column and a partially prestressed column (Sakai et.
al.2006).
Justification
The amount of post-tensioning was selected by keeping the initial force (10%f’cAg) in the post-tensioning at or below a tendon stress of 20% fpu. The amount of
longitudinal reinforcement that crosses the joint between the footing and the base of
the column was selected based on the literature review. Much of the previous posttensioning experiments had very little to no longitudinal reinforcement crossing the䋭282䋭joint between the footing and column as past researchers were just interested in the recentering effects of the tendon. Now that post-tensioning has shown excellent recentering effects, the combination of post-tensioning with longitudinal reinforcement
should be investigated. Sakai had a longitudinal reinforcement ratio of 0.65% with asingle tendon in his report. This amount of reinforcement allows for somelongitudinal column capacity in case of a tendon failure, and is small enough to allow
for minimal residual displacement. While the research described in this paper utilized
several tendons located around the center of the column cross section, it used a small
longitudinal reinforcement ratio to capture the re-centering effects for one column.
The second columns reinforcement ratio was doubled from the first to determine how
the amount of longitudinal reinforcement affects re-centering.
Design of Specimens and Test Setup
The two columns selected for testing were initially targeted to have identicalproperties except for the amount of longitudinal reinforcement crossing the joint
between the footing and the column base. Achieving identical forces in the tendons
between the two columns was difficult and ended with slight variations in the initialpost-tensioning (%f’cAg) between the two specimens. The column parameters areshown in Table 1 and the cross section for each column is shown in Figure 1. The
material properties for each column are shown in Tables 2 and 3.
The test setup for each column consisted of a reaction wall, strong floor, and a
220-kip actuator to produce the cyclic loading protocol in Figure 3. The reaction wall
was created by stacking two columns of three 4x4x8 foot (1.22x1.22x2.44 m)
concrete blocks, with one additional block added to the top of the front stack toachieve the proper height for the actuator to meet the column head. The reaction wall
was secured to the strong floor using 1.25 in. (31.75 mm) DywidagTM post-tensioned
bars. The footing of each column was secured to the 3 foot (0.91 m) deep strong floor
with six 1.25 in. (31.75 mm) DywidagTM post-tensioned bars. Each DywidagTM barfor the reaction wall and footing had a 100 kip (444.84 kN) force applied.
A steel spreader beam was bolted to the top of the column to apply the axial
dead load. Two hydraulic rams were attached to the top of the spreader beam on eachside of the column and two high strength 1.25 in. (31.75 mm) threaded rods ran
through the hydraulic ram, load cell, footing, and attached to the bottom of the strong
floor to apply the axial dead load. The axial dead load was maintained through an
accumulator which held constant and equal load in each of the hydraulic rams
throughout testing. The test setup is shown in Figure 4.

䋭283䋭

Results
The cyclic loading for each column produced a lateral force and displacement
hysteresis curve. The hysteresis curve was broken up into the positive backbone
envelope, considered the push cycle, and the negative backbone envelope, considered
the pull cycle. The absolute values from the negative backbone envelope weresuperimposed with the positive backbone envelope and an average curve was takenfrom the two envelopes. The average curve is considered the pushover curve for thecolumn. The pushover curve for columns PT-LL and PT-HL are shown in Figures 5
and 6 respectively.
Column PT-LL had a first bar yield displacement of 0.63 in (16.0 mm). Thiswas determined by finding the displacement when the first longitudinal bar yielded.
This value was used to plot a straight line on the pushover curve, beginning at zerodisplacement and zero force. A straight line was then plotted across the top of thecurve, where the area under the straight line bounded by the top of the pushover curve
was equal to the area under the pushover curve bounded by the straight line. The point
at which these two straight lines intersect is the effective yield displacement (0.95 in.
(24.2 mm). The ultimate displacement was defined as the displacement at 80% of thepeak lateral force in the column. The ultimate displacement of column PT-LL was
8.64 in (219.5 mm) at a drift of 8.0%, and a displacement ductility of 9.1. Column PTHL had a first yield displacement of 1.0 in (25.4 mm) that lead to an effective yielddisplacement of 1.36 in (34.5 mm). The ultimate displacement for column PT-HL was
9.17 in (232.9 mm) at a drift of 8.5% resulting in a displacement ductility of 6.7.
Each of the four tendons consisted of four 0.6” (15.24 mm) Grade 270 ksi(1862 MPa) 7-wire strands. Tendons 2 and 4 were located on the same axis that thecolumn was rotated about. Tendons 1 and 3 were located on the extreme ends and felt
the largest strains. The initial force in the tendons was carefully selected and kept
within the lower elastic region so the tendons would not yield under large lateraldisplacements. Figures 7 and 8 show the microstrain in tendons 1 and 4 with respect
to the drift ratio of the column. The outermost tendons (tendons 1 and 3) felt thelargest strains as shown in Figure 7, compared with tendons located on the same axis
the columns were rotated about (tendons 2 and 4) shown in Figure 8. The tendons do
not begin to yield until a microstrain of at least 8000 is reached. It can be seen thateven the most extreme tendons, such as tendon 1 did not reach their yield strains, evenat very large drift ratios. Column PT-HL behaved very similarly to column PT-LL anddid not reach a strain of more than 6500 microstrains, even at drift levels as high as
10%.

䋭284䋭

Figure 9 shows the full hysteresis curve for column PT-LL. During
construction, the cover concrete was almost double the amount on one side of thecolumn in relation to the other side. Uneven column cover lead to the core being
shifted and not in the true center, causing residual displacements and peak loads todiffer from a push and pull cycle. Figure 9 displays the enhanced re-centering effectprovided by post-tensioning. At a drift level of 6% (6.48 in, 164.6 mm), column PTLL had a residual displacement on the positive side of the hysteresis curve of 1.57 in(39.9 mm) and a residual displacement on the negative side of the hysteresis curve of
3.45 in (87.6 mm), resulting in an average residual displacement of 2.51 in (63.8 mm).
The full hysteresis for column PT-HL is shown in Figure 10. The residual
displacements are larger than PT-LL due to the increase in amount of longitudinal
reinforcement. At a drift level of 6% (6.48 in, 164.6 mm), column PT-HL had aresidual displacement on the positive side of the hysteresis curve of 3.12 in (79.2 mm)
and a residual displacement on the negative side of the hysteresis curve of 3.31 in(84.1 mm), resulting in an average residual displacement of 3.22 in (81.7 mm).
Conclusions
Two unbonded post-tensioned columns have been tested for seismic design.
The columns had identical properties except for the amount of longitudinal
reinforcement crossing the joint between the footing and the base of the column. Eachcolumn had four tendons evenly spaced around the center of the column cross section
to evenly distribute the force required for re-centering. While past research hasanchored the tendons in the base of the footing, this research anchored the tendons inthe side of the footing for the ease of inspection and replacement following anearthquake.
Tendon location and the initial tendon force selected (based on keeping thepost-tensioning at or below an initial tendon stress of 20% fpu) were satisfactory. The
extreme tendons (tendons 1 and 3) did not begin to yield, even at large drift ratios of
10% (10.8 in, 274.3 mm). Tendons located at 22.5% of the column diameter from thecenter of the column cross section provide re-centering capabilities and do not yield atlarge drift ratios. Exiting the tendons out the side of the footing did not display any
negative effects.
The amount of longitudinal reinforcement had a large impact on the recentering capabilities of the column. PT-LL had a residual displacement of 1.5 in(38.1 mm) at 6% drift (6.48 in, 164.6 mm), while column PT-HL had a residual
displacement of 3.1 in (78.7 mm) at 6% drift (6.48 in, 164.6 mm. The displacement
ductility of column PT-LL was 8.1 as opposed to 6.7 for column PT-HL.

䋭285䋭

Acknowledgments

Support for this research was provided by the Nevada Department ofTransportation. The findings herein are of the authors and do not necessarily represent
those of the sponsoring organization
Conversions
US Customery Units
1 in
1 ft
1 kip

SI Equivalent25.4 mm0.3048 m
4.45 kN.
References
Hewes, J. T., and Priestley, M. J. N. (2002). “Seismic Design and Performance ofPrecast Concrete Segmental Bridge Columns.” Structural Systems Research Project,
Rep. No. SSRP-2001/25, University of California, San Diego, La Jolla, California.
Ou, Y., Wang, P.-H., Tsai, M.-C., Chang, K.-C., and Lee, G. C. (2009). “Large-scale
experimental study of precast segmental unbonded post-tensioned concrete bridge
columns for seismic regions.” Journal of Structural Engineering, ASCE
Sakai, J., H. Jeong and S. Mahin, “Reinforced Concrete Bridge Columns that ReCenter Following Earthquakes,” Proceeding, 8th national Conference on Earthquake
Engineering, EERI, San Francisco, CA, April 2006, Paper 1421.
TABLE 1: COLUMN PARAMETERS
Columnȡlȡs

PT (initial)

Dead Load

0.685%
8%f'cAg, 157 kips 6%f'cAg
1.00%(10 #5's)(698 kN)(122 kips)
1.33%
9%f'cAg, 186 kips 6%f'cAg
PT-HL
1.00%
(10 #'7's)(827 kN)(122 kips)
PT-LL

Height, in(m)Diameter, in Aspect
(m)
Ratio

108 (2.74)

24 (0.61)

4.5

108 (2.74)

24 (0.61)

4.5

TABLE 2: CONCRETE PROPERTIESColumnSegment
7-Day Strength, psi (MPa) Test Day, psi (MPa)Footing4361 (30.1)
5384 (37.1)
PT-LL
Column & Column Head
3380 (23.3)
4330 (29.9)
Footing4361 (30.1)
5500 (37.9)
PT-HL
Column & Column Head
3380 (23.3)
4570 (31.5)

䋭286䋭

TABLE 3: STEEL PROPERTIES FOR COLUMN PT-LL
Tested Bar
fy, psi (MPa)fu, psi (MPa)Transverse Bars: #3
71300 (492)94500 (652)Longitudinal Bars: #5
71800 (495)96600 (666)Longitudinal Bars: #7
69800 (481)112200 (774)
Post-tensioned Strand: 0.6" 247000 (1703) 281000 (1937)

Tendon Duct
#3 Spiral 2" Pitch

Tendon Duct
#3 Spiral 2" Pitch
10 - #5 bars

(4)
(3)

10 - #7 bars

(4)

(1)

(1)

(3)
(2)

(2)
2'
Section A-A PT-LL

2'
Section A-A PT-HL(Tendon Numbers Shown in Parenthesis)

FIGURE 1: CROSS SECTION FOR PT-LL AND PT-HL

Note: Cross
section throughcenter line offooting.

Footing

FIGURE 2: TENDONS EXITING OUT THE SIDE OF THE FOOTING

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䋭299䋭

䋭300䋭SEISMIC PERFORMANCE ASSESSMENT OF CONCRETE BRIDGESDESIGNED BY DISPLACEMENT-BASED METHODS
Kevin Mackie1, Vinicio Suarez2, and Oh-Sung Kwon3
Abstract
With the advent of performance-based design, it is necessary to consider the
performance of bridges as an intrinsic part of the design process. However, even when
performance is measured in terms of deformations and displacement-based design isutilized, it is of interest to know whether designs actually result in the desired
performance under ground shaking representative of the design hazard. Four casestudies are designed in this paper, ranging from an elasto-plastic oscillator to a threespan continuous prestressed concrete bridge. The distribution of peak responses wasassessed for each case study in reference to the original target displacement used fordesign.
Introduction
Performance-based design (PBD) aims to improve performance by defining
performance criteria that must be satisfied at more than one earthquake level.
Generally, better performance is expected for important structures and smaller
earthquakes, while lower levels are required for ordinarily structures and more rare
events. Performance is no longer related to collapse prevention or life safety only;
deformations, functionality, economic losses, and downtime are additional criteria(Mackie and Stojadinovic, 2007; Mackie et al., 2010). Yet, there is a question as towhether the PBD procedure actually results in a structure that meets the performance
objectives.
PBD explicitly considers how a structure is likely to perform. The
performance assessment requires detailed analysis because it becomes an intrinsic partof the design process. A good preliminary design will reduce or eliminate the need for
iteration required to meet the performance objectives. Design procedures that are
useful within this framework must: (i) Take any combination of earthquake level andperformance criteria; (ii) Produce a design that meets the target performance; (iii) Be
rational and easy to execute. Both force-based and displacement-based procedures arepotential design methodologies for applicability as PBD tools; however, each withdiffering merits in terms of the three criteria listed above.
Regardless of the design methodology adopted; however, design should be
carried out considering the following. First, seismic resistant bridges should have a
1

Assistant Professor, Department of Civil, Environmental, and Construction
Engineering, University of Central Florida, Orlando FL 32816-2450, USA.
2
Director UCG, Universidad Tecnica Particular de Loja, Loja, Ecuador.
3
Assistant Professor, Department of Civil Engineering, University of Toronto,
Toronto ON M5S 1A4, Canada.

䋭301䋭simple configuration, such that their behavior can be easily modeled and analyzed.
The chosen configuration should also aim to include energy dissipation in different
components of the structure with ductile mechanisms. Second, in conventional
bridges, pier columns provide the primary energy dissipation mechanism, whileabutments can provide additional energy dissipation (Priestley et al., 1993).
Recommended earthquake resisting systems for bridges are given in the Guide
Specifications for LRFD Seismic Bridge Design (AASHTO, 2009). Finally, capacity
design principles must be applied in all cases to protect the components outside theductile mechanism and to prevent non-ductile modes such as shear.
Force-Based Design (FBD) and Displacement-Based Design (DBD)
Seismic design of bridges can be accomplished following different
approaches. The traditional procedure is force based since damage in the structure iscontrolled by the assignment of a certain level of strength. The procedure uses
strength reduction factors to reduce the elastic force demand while considering
importance, assumed ductility capacity, over-strength and redundancy in the structure.
FBD is found in the AASHTO LRFD Bridge Design Specification (AASHTO, 2004)
and was first adopted by AASHTO in 1983 following recommendations of the
Applied Technology Council (ATC, 1981).
There are several problems attributed to FBD. First, strength is used as ameans to control damage, although these parameters do not correlate well. Second, it
is assumed that strength and stiffness are independent. Third, force reduction factors(R) are used assuming that the ductility demand will be the same for each type ofstructure. Finally, the R factors are given generally for a single level “no-collapse”
design. Multi-level design would require the specification of different R values.
After the Loma Prieta earthquake in 1989, extensive research has been
conducted to develop improved seismic design criteria for bridges, emphasizing theuse of displacements rather than forces as a measure of earthquake demand anddamage in the structure. Research has also focused on the application of capacity
design principles to assure ductile mechanisms and concentration of damage in
specified regions. Several DBD methodologies have been developed including:• Direct Displacement Based Design (DDBD) (Priestley, 1993)• MCEER/ATC-49 Recommended LRFD guidelines for seismic design ofbridges (ATC, 2003)• Seismic Design Criteria (SDC) of Caltrans (Caltrans, 2004)• Guide Specifications for LRFD Seismic Bridge Design (AASHTO, 2009)
DBD has gained popularity in the last fifteen years, as it addresses several
shortcomings of the conventional FBD procedure, while serving as a useful tool for
performance-based seismic design. The primary difference between DBD and FBD isthat the former uses displacement as a measure of seismic demand and also as anindicator of damage in the structure. DBD takes advantage of the fact that
displacement correlates better with damage than force. DBD also overcomes serious
problems of FBD such as ignoring the proportionality between strength and stiffness䋭302䋭and the generalization of ductility capacity through the use of force reduction factors.
DBD can be used with any combination of earthquake level and performance criteria.
Both conventional DBD and direct DBD are compared briefly below;
however, DDBD is selected as the design procedure for the case studies presented inthis paper. The primary differences in the AASHTO (2009) LRFD procedure and
DDBD are linearization and the execution. The displacement demand assessment
procedure in the AASHTO LRFD Seismic guide uses elastic analysis and the equal
displacement approximation (Veletsos and Newmark, 1960) to obtain inelastic
displacement demands (an amplification factor is used with short period structures).
In the elastic analysis the structure is modeled with cracked section stiffness. DDBD
uses the equivalent linearization to overcome the limitations of the equal displacement
approximation (Suarez, 2008). In execution, the AASHTO LRFD uses a
demand/capacity assessment procedure. In contrast to this, DDBD goes directly fromtarget performance to required strength. The amount of reinforcement does not need
to be assumed during design.
Conventional Displacement-Based Design approach
The Seismic Design Criteria by Caltrans (2004) shifted towards displacementbased design in 1999 by consolidating ATC-32 recommendations (ATC, 1996).
Currently, Caltrans has an iterative design procedure in which the lateral strength of
the system (size and reinforcement of the substructure sections) is assumed at the
beginning of the process. Then, by means of displacement demand analysis anddisplacement capacity verification, it is confirmed that the bridge has an acceptableperformance, otherwise, the strength is revised and the process repeated.
In the demand analysis, the peak inelastic displacement demands are estimatedfrom a linear elastic response spectrum analysis of the bridge, with cracked (secant toyield point) component stiffness. Then, elastic peak displacements are converted topeak inelastic displacements using an equal displacement approximation (Veletsos
and Newmark, 1960) with modification for short period structures. Once thedisplacement demands are estimated, the procedure requires the verification of thedisplacement capacity of each pier by means of a pushover analysis. Finally, thesubstructure sections and protected elements are designed and detailed according tocapacity design principles.
The AASHTO Guide Specifications for LRFD Seismic Bridge Design (2009)
recognizes the variability of seismic hazard over the US territory and it specifies
different Seismic Design Categories (SDC). Each SDC links seismic hazard to
expected performance. The design procedure in the AASHTO DBD approach is inconcept similar to the Caltrans approach. Depending on the configuration of thebridge, the demand analysis is performed by the uniform load method for regularbridges, while spectral modal analysis can be used for all bridges. The capacity
verification can be done using implicit equations for seismic design category B or bypushover analysis for categories C and D. As with the Caltrans approach, and withthe exception of seismic category A, the proposed guide requires the use of capacity

䋭303䋭design principles for the detailing of the substructure sections and protected elements.
Current Caltrans and AASHTO approaches utilize acceleration spectrum
curves to determine the displacement demand at the system level. The main
limitations of this approach are:
• The use of the equal displacement approximation. Research conducted on
displacement modification factors (FEMA 440, 2005) has shown that the ratio
between inelastic and elastic displacement depends on period, hysteresis shape, andother factors. In addition to this, assuming that the elastic displacement demand
equals the inelastic demand is not appropriate when additional damping exists in the
structure as a result of soil-structure interaction or other sources.
• The use of acceleration response spectrum to compute displacement demand.
The displacement spectrum seems a more rational source of seismic hazard for DBD.
• The procedure is iterative in nature since reinforcement in the pier sections
must be guessed at the beginning of design. If the displacement capacity is ultimately
less than the displacement demand, the process must be repeated increasing theamount of reinforcement. If the inverse occurs, no iteration is needed; however, theresulting design will be overly conservative.
Direct Displacement Based Design (DDBD)
DDBD has been conceived as a tool to achieve deterministic PBD, as a simple
methodology that can be used to go from basic geometry to properly detailed sections
and structural components. Research conducted in the last fifteen years have shown
the method produces satisfactory bridge designs (Kowalsky et al., 1995, Calvi andKingsley, 1995, Dwairi et al., 2006, Suarez and Kowalsky, 2007, Priestley et al.,2007), however a formal reliability study has not been yet conducted and the use of
DDBD within the scope of probabilistic PBD requires further research. The DDBD
method was initially proposed by Priestley (1993). In its current state, DDBD works
with any combination of seismic hazard and performance criteria, and it is intended to
produces structures that meet (theoretically in the mean), rather than be bounded by,
the target performance. This makes DDBD a very attractive alternative forpreliminary design since it reduces, and in some case might eliminate, the need for
iteration in a general PBD procedure.
DDBD differs from the DBD procedure in the AASHTO Guide Specifications
for LRFD Seismic Bridge Design in the use of an equivalent linearization approachand in the execution of the procedure. DDBD starts with the definition of aperformance-based target displacement for the structure and returns strength required
to meet the target displacement under the specified earthquake. The method isreferred to as “direct” since, in contrast to the traditional DBD procedure of AASHTO
or Caltrans, the reinforcement and thus the strength of the structure does not need to
be assumed at the beginning of the design and modified iteratively until ademand/capacity check is satisfied.

䋭304䋭

(a) Equivalent Linearization(b) Displacement Spectrum

FIGURE 1 - Fundamentals of DDBD
DDBD uses an equivalent linearization approach (Shibata and Sozen, 1976)
where an inelastic system at maximum response is modeled by an equivalent elastic
system with secant stiffness (Keff) and equivalent viscous damping (ξeq) (FIGURE 1a).
A design objective must be defined as a combination of a performance criterion anddesign earthquake. The performance can be specified in terms of material strains,
curvature, drift, or ductility in the piers. In all cases, consideration of abutment
displacements as a limit state, as well as P-Δ effects of bridge piers should also be
addressed. The design objectives can also be those stated for each SDC in the
AASHTO LRFD Seismic Guide Specification. The design earthquake is represented
by a displacement spectrum that is reduced to the level of damping of the structure(FIGURE 1b).
In most cases DDBD can be applied with simple hand calculations. Modal
spectral analysis or pushover analysis are not required. A major limitation of thisapproach is the target displacement can only be estimated for simple pierconfigurations (the most common at least). Since a pushover analysis is not carriedout, the flexibility of cap beams and rotation of foundations (for example) cannot be
incorporated without some iteration. In addition, the procedure is direct (no iteration)
only when the shape of the displacement profile is known. This scenario occurs onlyin bridges with regular distributions of mass and stiffness. Curved bridges are design
as straight.
Analysis Procedure
A DDBD procedure is carried out for four bridge case studies in this paper. A
location is selected in the central United States to define an equal hazard spectrum for
all four case studies, except for the final bridge structure that has a higher design
spectrum. After a target displacement is selected for each structure based on
appropriate limit state definitions (specific to reinforced concrete bridges), theperformance of the case study structures was assessed using nonlinear time history
analysis (THA) with recorded ground motions. The probabilistic assessment allows
confirmation of whether the target displacement is achieved in the mean, reflecting an
unbiased design procedure. Due to the assumptions surrounding the DDBD
procedure, meeting the performance objective in the mean is unlikely; therefore, this䋭305䋭paper investigates whether the design procedure can be formulated to achieve a
conservative design. Above and beyond this, it is demonstrated that under a given
performance objective and probabilistic acceptance criterion, a modifier on the initialdesign displacement can be specified so that the procedure retains the advantages ofbeing non-iterative (see FIGURE 2).

FIGURE 2 - Probabilistic considerations in DDBD and assessment
A complicating factor for nonlinear assessment of structures designed
according to DDBD is that bridges are inherently three-dimensional (3D) systems.
Two important phenomena are illustrated in the case studies presented in this paper:
1.) 3D excitation and response impact performance in ways that are difficult toaccount for in only longitudinal or transverse simplifications (or combination rules),and 2.) the bridge responds as a system, with contributions from several load-resisting
components such as shear keys, abutments, foundations, and the superstructure itself.
Therefore, assessment often yields different response quantities than the initial design.
The choice of assessment procedure and degree of model complexity also influenceagreement between design and assessment. Allowance for different assessment
techniques (linear dynamic, nonlinear static, nonlinear dynamic, etc.) should beconsidered when proposing any modification factors on the initial target
displacement; however, only nonlinear THA is considered in this paper.
Hazard, DDBD, and Case Study Details
A site near the New Madrid seismic zone was selected (-90.196, 35.212).
Multi-level hazards are defined based on USGS seismic hazard maps. Spectral
acceleration curves were generated for three hazard levels: 10%-, 5%-, and 2%probability of exceedance in 50 years and 5% equivalent viscous damping. The
acceleration spectra were converted to displacement spectra and then linearized(relationship between T and Sd). The peak spectral displacement was assumed tooccur at the corner period of 3 sec. Based on extrapolation of the Sd curve, the Sd at
corner period are 5.76, 16.1, 33.4 (cm) for each of the three hazard levels. The
spectral displacement hazard at the 2% probability of exceedance in 50 year level was䋭306䋭raised to 72 cm for the bridge in case study as it was originally designed for a site in
California with substantially higher hazard.
Once the design objective has been selected, the main steps of the DDBD
procedure are: (i) Select the dimensions of the components of the earthquake resisting
system on the basis of past experience. (ii) Compute a target displacement based on
the performance level for the structure. Depending on whether this is specified as astrain, ductility, etc., it may be necessary to use a plastic hinge model to relate these to
peak displacements. (iii) Evaluate the effective mass and equivalent viscous damping
for the system. Priestley et al. (2007) defined a ductility vs equivalent damping
relationship. (iv) Compute the spectral reduction factor that corresponds to theequivalent damping level in the structure and find a reduced design spectrum. The
spectral reduction factor for inelastic structures is based on Eurocode (1998). (v)Determine the required effective period, secant stiffness and required strength. (vi)Distribute the required strength, design plastic hinges and protected elements using
capacity design principles.
The four case studies considered in this paper are: 1.) a single-degree-offreedom elasto-plastic oscillator, 2.) a two-dimensional reinforced concrete bridge
bent with a single column subject to transverse excitation, 3.) the same twodimensional bridge bent subject to both lateral and vertical excitation, and 4.) a 3D
reinforced concrete bridge with 3 continuous spans and explicit foundation andabutment representations.
Case Study 1: Single-degree-of-freedom (SDOF) system
The SDOF system was selected as the simplest case of design where theperiod of the structure (both loading and unloading) is constant and the yield point
defines the perfectly plastic plateau. Therefore, the only unknowns are the period andthe yield strength. A factor of 55 was used in the expression for equivalent dampingof the elasto-plastic system. The benefit of using a SDOF oscillator is that it is notnecessary to use nonlinear THA to assess the performance of the system, otherapproximate techniques can be readily used (such as R--T relationships). The
oscillator was assumed to have a yield displacement of 0.05 m (typically the yielddisplacement is obtainable from the structure’s geometry) and a target ductility of 4.
For the 2%-in-50-year hazard defined previously, the effective period becomes 3.28
sec and the target strength is 0.73 kN.
The properties of the oscillator were then used in two separate analyses. The
first was to use a common relationship between R--T to obtain the achieved ductility
at the target spectral displacement. The second was to perform inelastic THA with thespecified oscillator properties. A total of 160 ground motions were scaled to the target
spectral displacement before performing the analysis and the distribution of achieved
maximum displacements is shown in FIGURE 3 below. The individual groundmotion realizations are shown in the top pane of the figure while the cumulative
distribution function (CDF) is shown in the bottom pane. The actual time history CDF
is shown with lower (LCB) and upper (UCB) bound indicators. Finally, the data is

䋭307䋭The two bents are skewed 20 degrees and have two 1.83 m diameter columns
supported on piles (FIGURE 8). Column height varies from 13.4 m at bent two to
14.3 m at bent three. The columns are pinned at the bottom and fixed to an integral
bent in the superstructure. The bridge is founded on seat type abutments withelastomeric bearings and a break-off wall once the gap closes in the longitudinal
direction. Exterior shear keys prevent transverse motion under lower intensity
motions and service loads. The superstructure is capacity designed to remain elastic at
the target displacement and is therefore modeled using an elastic section with cracked
properties. No explicit representation of the tendons or mild steel was created in theanalytical model.

FIGURE 7 - Elevation of 3-span continuous case study bridge (Suarez and Kowalsky,
2010)

FIGURE 8 - Bent configuration and superstructure cross section for 3-spancontinuous case study bridge (Suarez and Kowalsky, 2010)
The expected concrete compression strength was 36000 kPa, expected yieldstrength of main reinforcement 455000 kPa, diameter of spiral 25 mm, pitch of spiral
125 mm, and 50 mm cover to main reinforcement were consistent with the DDBD
design. The superstructure elastic properties were obtained from the LRFD design
example appendix and were factored by 0.5 for cracked moment of inertia and 0.25
for cracked torsional constant. The bent cap is modeled explicitly and also contains
cracked elastic properties based on initial gross dimensions. The two columns per䋭311䋭is 50% (most likely too high from a risk perspective), then the inverse problem can besolved. A modifier on the target displacement (0.28 m) will result in a new period andrequired strength (0.523 kN). Assessment of this modified system yields a mean
displacement of 0.2 m. For the SDOF system, this phenomenon is easily exploredanalytically using the expression for spectral factor based on the equivalent damping.
Continuing work will demonstrate the relationship between the distribution of thisparameter and the resulting responses. The numerical level of conservatism (andtherefore the ability to achieve target risk levels for each displacement) will bedemonstrated and related to the brief inverse problem described in this section. It isalso worth noting that as the target displacement (with modifier) is increased, theselection of recorded ground motions that meet the target spectral displacement for
design diminish. Therefore, scale factors on ground motion amplitude are used thatwould likely cause a bias in the observed response distribution.
Conclusions
This paper investigates performance-based assessment of the direct
displacement-based design (DDBD) procedure. Four case studies were selected
ranging from an elasto-plastic oscillator to a three-span continuous prestressed
concrete box girder bridge. Each case study was designed using DDBD for a specified
hazard level. Subsequently, a nonlinear time history analysis was conducted to assess
the performance of each, or more specifically, the probabilistic distribution of peak
displacement responses in reference to the original target displacement used fordesign. It is demonstrated that, consistent with earlier findings, the DDBD leads to aslightly conservative design whereby the target displacement is exceeded less than
50% of the time. In all the case studies (except Case 3), the probability of exceeding
the target displacement is approximately 25%. It was demonstrated that thisinformation enables a modifier to the original target displacement to achieve a
specified risk level (acceptable probability of exceeding target displacement) without
iteration. Further work is necessary to determine the nature of the response
distribution for different structure types and modeling assumptions before such atechnique can be used more broadly to achieve performance-based design objectives.
References
AASHTO, 2006, LRFD Design Example B November 3, 2006 – Version 1.1,
http://cms.transportation.org/?siteid=34&pageid=1800, (accessed April 18, 2008).
Berry, M.P., and Eberhard, M.O., 2003, Performance models for flexural
damage in reinforced concrete columns, Report No. 2003/18, Pacific Earthquake
Engineering Research Center, University of California Berkeley.
Calvi G.M. and Kingsley G.R., 1995, Displacement based seismic design ofmulti-degree-of-freedom bridge structures, Earthquake Engineering and Structural
Dynamics, 24, 1247-1266.
Dwairi, H. and Kowalsky, M.J., 2006, Implementation of inelastic
displacement patterns in direct displacement-based design of continuous bridge
structures, Earthquake Spectra, 22(3), 631-662.
EuroCode 8, 1998, Structures in seismic regions – Design. Part 1, General and䋭313䋭Building”, Commission of European Communities, Report EUR 8849 EN
Kowalsky M.J., Priestley M.J.N. and MacRae G.A. 1995. Displacement-based
design of RC bridge columns in seismic regions, Earthquake Engineering andStructural Dynamics, 24, 1623-1643.
Mackie, K.R., and Stojadinovic, B. 2007. Performance-based seismic bridge
design for damage and loss limit states. Earthquake Engineering and Structural
Dynamics, 36(13), 1953-1971.
Mackie, K.R., Wong, J-M., and Stojadinovic, B. 2010. Post-earthquake bridge
repair cost and repair time estimation methodology. Earthquake Engineering andStructural Dynamics, 39(3), 281-301.
Priestley, M. J. N., 1993, Myths and fallacies in earthquake engineeringconflicts between design and reality, Bulletin of the New Zealand Society ofEarthquake Engineering, 26 (3), 329–341.
Priestley, M. J. N., Calvi, G. M. and Kowalsky, M. J., 2007, Direct
Displacement-Based Seismic Design of Structures, Pavia, IUSS Press.
Shibata A. and Sozen M. Substitute structure method for seismic design in
R/C. Journal of the Structural Division, ASCE 1976; 102(ST1): 1-18.
Suarez, V.A. and Kowalsky, M.J., 2007, Displacement-based seismic designof drilled shaft bents with soil-structure interaction, Journal of Earthquake
Engineering, 11(6): 1010 – 1030.
Suarez, V.A., 2008, Implementation of direct displacement-based design for
highway bridges, PhD Dissertation, North Carolina State University.
Suarez, V.A. and Kowalsky, M.J., 2010, Direct Displacement-Based Design
as an Alternative Method for Seismic Design of Bridges, ACI Special Publication,
SP-271(4): 63 - 78.
Veletsos, A. and Newmark, N. M., 1960, Effect of inelastic behavior on the
response of simple systems to earthquake motions”, Proceedings of 2nd World
Conference on Earthquake Engineering, pp. 895 – 912.

䋭314䋭

LOCAL BUCKLING ANALYSIS OF STEEL TRUSS BRIDGE UNDERSEISMIC LOADING
Eiki Yamaguchi1, Keita Yamada2
Abstract
In the 2004 Mid Niigata Prefecture Earthquake, a steel truss bridge wasdamaged: the lower chord member underwent local buckling. The axial force in thatmember is not necessarily compression-dominant: tensile axial force is also expected.
Since many steel bridge piers were subjected to local buckling in the 1995 Kobe
Earthquake, the criterion for local bucking in the member under axial compression has
been studied rather extensively. However, the local buckling in the member under theother states of axial force has not. In the present study, the local buckling in the lower
chord member of a truss bridge is to be looked into. To that end, the existing criterion
for local buckling in terms of average strain is tested for the case when tensile yielding
precedes compression, failing to confirm its applicability. Then the criterion ismodified by introducing the updated average strain. The seismic response analysis isthen conducted to show the significance of the proposed criterion.
Introduction
One of the largest earthquakes in the recorded history, the Tohoku Earthquake,just hit Japan in March, 2011, causing very serious damage in the eastern part of Japan.
Yet the memory of the damage to structures in the 1995 Kobe Earthquake is still freshand vivid for many structural engineers. Between the two large earthquakes, numerous
earthquakes occurred as well, some of which were quite large and comparable to the1995 Kobe Earthquake. The damage in each big earthquake has posed a new challenge
for engineers; some of them are yet to be solved.
In the 2004 Mid Niigata Prefecture Earthquake, a steel truss bridge wasdamaged: the lower chord member underwent local buckling at its fixed end. The axial
force in that member is not necessarily compression-dominant: tensile force can be
expected. Since many steel bridge piers experienced local buckling in the 1995 Kobe
Earthquake, the criterion for local bucking in the member under axial compression has
been studied rather extensively (Ono et al. 2007, Committee 2008). However, the local
buckling of the member under the other states of axial force has not. In the present
study, the local buckling in the lower chord member of a truss bridge is to be looked
into.
Existing Criterion for Local Buckling
Local buckling can be simulated in the finite element analysis (FEA) with shell
1

Professor, Dept. of Civil Engineering, Kyushu Institute of Technology, Kitakyushu
Graduate Student, Dept. of Civil Engineering, Kyushu Institute of Technology,
Kitakyushu
2

䋭315䋭

elements. Even though it is not impossible to model the whole bridge by shell elements
and conduct nonlinear dynamic FEA, that is not a practical approach to structuraldesign; beam elements are employed exclusively for the analysis of seismic design.
However, since the cross section of a beam element does not deform, the direct
simulation of the local buckling by beam elements is not possible. To overcome thedifficulty, various efforts have been made, which includes the detection of the local
buckling by the magnitude of strain (Ono et al. 2007, Committee 2008) and theimplementation of strength reduction due to local buckling in the constitutive
relationship (Yamaguchi 2009).
The criterion of the local buckling due to Committee (2008) is based on the
average strain İ in a compressive flange over the characteristic length LC. For abox-section member, the characteristic length LC is given by
LC

Min(0.7b, a)(1)

where a is the distance between two adjacent diaphragms and b the width of a flange.
Min indicates that the smaller of the two values in the parenthesis shall be taken. On theother hand, the limit strain İu for an unstiffened box-section member is computed by
Hu
Hy

0.24
R f  0.2

2 .8

1  Nc N y

2.4



2.8
1  N c N y 0.6

d 20.0(2)

where İy is the yield strain, Rf the width-to-thickness ratio parameter, NC thecompressive axial load and Ny the squash load. The validity of the equation has beenverified for 0.2 d Rf d 0.7㸪0.0 d N/Ny d 1.0.
The criterion for the local buckling is then that the member is judged to undergo
local buckling at the instance when the average strain İ reaches the limit strain İu. It isnoted that NC changes during earthquake so that the limit strain İu varies with time as
well.
Equation 2 has been obtained under compressive loading appliedmonotonically. Therefore, the validity of Equation 2 is not clear if tensile yielding
precedes compression. The investigation is needed herein since the lower chordmember in a truss bridge may yield in tension.
To this end, a short box-section member shown in Figure 1 is constructed. It ispulled first and then compressed until local buckling occurs. This nonlinear problem isanalyzed by ABAQUS (Dassault 2008) using 1280 shell elements: the local buckling
can be simulated directly. The material is steel with Young's modulus E equal to2.0x105 N/mm2 and the yield stress ıy equal to 235 N/mm2. The stress-strain
relationship is of a bilinear type with the slope after yielding being E/100 (Figure 2).
Six cases are considered, between which the difference lies in the initial䋭316䋭1200
800
400
0
-400
-800
-1200
0

5

10

15
20
TIME (sec)25

30

Figure 8 Acceleration recorded in 2004 Mid Niigata Prefecture Earthquake
The updated average strain İ' is plotted in Figure 9 (b). It is very different fromthe average strain İ in Figure 9 (a): the updated average strain İ' does meet the limit
strain İu so that the local buckling is judged to take place. The significance of the
proposed criterion is thus obvious.
Concluding Remarks
When beam elements are employed for the seismic response analysis of abridge, the criterion for local buckling is necessary. However, it has been concluded in
the present study that when tensile yielding preceded, the existing criterion could notgive the correct judgment.
The modification of the criterion in which the average strain was replaced by
the updated average strain has been proposed. The seismic response analysis of a steel
truss bridge then demonstrated the significance of the proposed criterion.
References
Chen, W.F.: Constitutive Equations for Engineering Materials, Vol.2. Amsterdam,
Elsevier, 1994.
Committee on Steel Structures: Standard Specifications for Steel and Composite
Structures - 2008 IV Seismic Design, Tokyo, Japan Society of Civil Engineers (JSCE),
2008.
Dassault Systemes Simulia Corp.: ABAQUS: User's Manual, ABAQUS Ver. 6.8,
2008.
Ono, K., Hashimoto, A. Nishimura, N., Yamaguchi, E.: Seismic evaluation method of
stiffened box members with fiber model, Bridge and Foundation Engineering, Vol. 41,
No. 6, pp. 26-33, 2007.
Yamato Design Co., Ltd.: User's Manual for Y-FIBER3D, 2000.

䋭321䋭

27th US-Japan
Bridge Engineering Workshop

Session8

Construction and Maintenance3
Use of Precast I-Girders for Accelerated Bridge Construction in High Seismic Regions
By Sri Sritharan, Justin Vander Werff, and Rick Snyder
How to Build a Bridge - Fast
By John Stanton
Research and Repair of Reinforcement by Overlaying Floor Slab Method in Expressway
By Hidekazu Hayashi
Repairing Fatigue Cracks in a Steel Deck Plate Girder Bridge of Unusual Structure
By Haruhiko Nakata, Yoshiyuki Takamura, and Ken Tokumasu
Development of Eddy Current Test to Detect Fatigue Damages on Orthotropic Steel Deck
By Hiroki Sugiyama, Akiko Tabata, Hiroyuki Nakajima, Tetsuji Yamagami,
Sigeaki Tsukamoto, Akira Shiraishi, and Takaaki Yamada

Page Left Blank

䋭324䋭

USE OF PRECAST I-GIRDERS FOR ACCELERATED BRIDGECONSTRUCTION IN HIGH SEISMIC REGIONS
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AASHTO LRFD Bridge Design Specifications, 3rd Edition, $$6+72 
Accelerated Bridge Construction Applications in California, A Lessons Learned Report,
Version 1.1, &DOWUDQV 
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IRU 6WDQGDUG %ULGJHV´ PCI Journal, 9  1R  -XO\$XJXVW  SS 
Bridge Design Aids, &DOWUDQV 
Bridge Design Specifications, &DOWUDQV 
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3RVLWLYH 0RPHQW &RQQHFWLRQV LQ 3UHVWUHVVHG &RQFUHWH *LUGHUV 0DGH&RQWLQXRXV´ Journal of Bridge Engineering,  SS 
3ULHVWOH\ 01 6HLEOH ) DQG 8DQJ & The Northridge Earthquake of January 17,
1994, 7KH 8QLYHUVLW\ RI &DOLIRUQLD 
Seismic Design Criteria, Version 1.4, &DOWUDQV 
6Q\GHU 5LFKDUG 0 Seismic Performance of an I-girder to Inverted-T Bent Cap Bridge
Connection, 
6Q\GHU 5LFKDUG 0 9DQGHU :HUII - 7KHLPDQQ =DFKDU\ - 6ULWKDUDQ 6UL DQG+RORPER -D\ Seismic Performance of an I-Girder to Inverted-T Bent Cap
Connection )LQDO &DOLIRUQLD 'HSDUWPHQW RI 7UDQVSRUWDWLRQ ,RZD 6WDWH
8QLYHUVLW\ 
7KHLPDQQ =DFKDU\ - 3-D Finite Element Analysis of the Girder-to-Cap Beam
Connection of an Inverted-tee Cap Beam Designed for Seismic Loadings, ,RZD
6WDWH 8QLYHUVLW\ 
WinRECOL, 75&,PEVHQ 6RIWZDUH 6\VWHPV
Xtract, 75&,PEVHQ 6RIWZDUH 6\VWHPV

䋭335䋭

䋭336䋭HOW TO BUILD A BRIDGE - FASTJohn Stanton1
Abstract
A new connection has been developed for connecting precast bridge columns to castin-place spread footings. It is quick and simple to construct, and has excellent seismic
resistance as well. It is made by precasting the column; setting, plumbing and levelingit on site; fixing the footing steel; and casting the concrete for the footing. No steel
crosses the interface between the precast column and the spread footing; the column
bars are all straight and are terminated with heads for anchorage. The column has aroughened surface that transfers the shear stresses (due to column axial load andbending) across the column-footing interface. Tests were carried out on the system.
The paper describes the design criteria and methodology, the results of the tests, andrecommendations for use of the system in practice.
Introduction
Bridge construction frequently leads to traffic delays, which result in wasted
time and fuel. Bridge owners are therefore seeking methods to accelerate bridge
construction, referred to as ABC. Such methods also offer reduced environmental
impacts, better worker safety, higher quality construction and lower lifecycle costs(Wacker et al. 2005). Use of precast concrete represents a promising technology for
ABC, and has been successfully used for bridge substructures in non-seismic regions(Matsumoto et al. 2001). Connections are typically made at the beam-column andcolumn-foundation interfaces to facilitate fabrication and transportation. However, forstructures in seismic regions, those interfaces represent the locations of high moments
and large inelastic cyclic strain reversals. Devising connections that are not only
sufficiently robust to accommodate those inelastic cyclic loads, but are also readily
constructible, is challenging.
A bridge bent system has been developed at the University of Washington, in
collaboration with WSDOT and Berger/ABAM Engineers, that is intended to satisfy
the combined needs of seismic performance and rapid construction. Different
connection systems are used at the column-to-foundation and the cap beam-to-column
interfaces, because of the different conditions that exist at each location. The capbeam connection is described in Pang et al. (2010) and is shown in Figure 1. It
consists of bars that project from a precast column and are grouted into ducts in the
precast cap beam. The distinction between it and other grouted bar systems is thatlarge bars (up to #18 US, or D57) are used in large ducts. That approach allows theuse of a small number of bars, which minimizes the number of bar fit-ups needed onsite and maximizes the size of the ducts, both of which improve constructability.

1

Professor, University of Washington, Seattle, WA 98195, USA.

䋭337䋭

FIGURE 1. COLUMN TO CAP BEAM

FIGURE 2. CONSTRUCTIONCONNECTION CONCEPT. (1 inch = 25 mm)

SEQUENCE

The potential drawback of the use of large bars is their anchorage within thedepth of the cap beam. However, extensive testing (Steuck et al. 2009) showed thateven #18 (D57) bars could easily be anchored within the depth of a typical cap beamand that, under cyclic lateral load, the connection behaved like a cast-in-place systemwith conventional detailing and bar sizes (Pang et al. 2010).
This paper describes the socket connection between the column and spreadfooting, for which the same requirements, namely simplicity of construction and goodseismic performance, exist. Research to extend the concept to the use of drilled shaftsis ongoing.
The construction sequence for the new socket connection is shown in Figure
2. The column is precast with a roughened outer surface at the bottom. Once thefooting has been excavated (Step 1 in Figure 2), the precast column is brought to site,plumbed, leveled, and braced (Step 2). Footing reinforcement is then placed, and thefooting is cast (Step 3) around the column.
The final step is to connect the column to the precast cap-beam (Step 4) bygrouting the large bars in the large ducts. In comparison with conventional cast-inplace construction, the primary advantage of this system is construction speed; afooting and column can be built in little more time than is needed to cast the footing
alone. Further, the use of a precast cap beam is estimated to save several weeks.
The structural details differ from those of a conventional, cast-in-place system
in two ways. First, no bars pass from the footing into the column, so the only

䋭338䋭Guide Specifications. Therefore they were included too. The primary requirementsare:•The soil pressure under the footing, under vertical load plus overturning, mustremain below the allowable bearing pressure.

•

The resultant vertical reaction, under vertical load plus overturning, mustremain within the middle two-thirds of the footing.

•

The connection between the footing and column must provide sufficientstrength to force the failure to occur in the column. The goal is to avoid
footing failure, which can be expensive to inspect and repair.

The first two requirements, based on soil properties, essentially define the plan
dimensions of the footing and are not discussed further here. The structuralrequirements for the connection can be identified from the potential modes of failure.
The primary ones are:•

Bending strength of the footing.

•

One-way shear strength of the footing.

•Anchorage failure of the bars.

•

Punching shear failure under vertical load (assumed to occur on a conicalsurface).

•

Transfer of combined vertical load and moment between the column andfooting.

•

Shear friction failure across the precast to cast-in-place interface.

•

Joint-shear failure within the connection.

The longitudinal column bars are equipped with headed anchors, which should beselected from commercially available products and may thus be assumed to provide
adequate anchorage. The other structural requirements depend on the flow of forces
in a system that is highly statically indeterminate. While the distribution of internal
forces is expected to follow the general pattern shown in Figure 3, the resistance ineach potential failure mode is not well defined and so needs to be determined by
testing. The characteristics that were expected to influence the resistance were:•

The ratio of footing depth to column diameter.

•

The quantity of column longitudinal reinforcement.

•

The quantity of shear reinforcement in the column within the joint region.

•

The quantity of transverse reinforcement in the footing.

Test Program
The number of characteristics exceeded the number of tests (three) that could
be conducted within the program resources, so choices had to be made. All the test
specimens consisted of cantilever columns projecting from spread footings, in which
the columns were subjected to constant vertical load and cyclic lateral load. The test
specimens consisted of 20-in. (500 mm) diameter columns and represented at 5/12

䋭340䋭

scale a notional prototype with a 4-ft (1200 mm) diameter column. Specimens SF-1
and SF-2 each contained a column splice above the plastic hinge region. However,the splice is not an essential part of the system, and is not discussed further here.
In many columns in the field, the longitudinal column reinforcement ratio is
close to the AASHTO minimum of 1%. Therefore this ratio was used for all threetests. The spiral steel in the column is also typically carried down into the footing at
the same pitch as in the body of the column, so that was also done in all of these test
specimens. It provided joint shear some resistance.
The primary test variables were the ratio of footing thickness to column
diameter and the quantity of transverse steel used in the footings. In addition several
other details of the footing steel were varied between specimens. The specimen
details are summarized in Table 1. The test program was carried out while theWashington State Department of Transportation, WSDOT, was designing a bridgethat uses the system. The bridge construction started only eight months after thatlaboratory testing was complete, and is now finished. The geometry of the test
specimens represented, at 1: 2.4 scale, the geometry of the prototype bridgecomponents.
TABLE 1. TEST SPECIMEN DETAILS
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(Note: 1 inch = 25 mm).
Specimen SF-1 was regarded as the most conservative detailing, intended torepresent as closely as possible a direct conversion from cast-in-place to precast
construction. Thus some of the flexural steel in the bottom of the footing passed
directly under the column, and this required casting a slot into the bottom of thecolumn, as shown in Figure 4. The goal was to ensure the best possible engagementof the footing steel with the column steel. A plan view of the test specimen is also
shown in Figure 5.
In all three specimens, the column projected slightly below the structural partof the footing in order to locate the anchor head on the column bar just below thebottom node in the strut and tie model shown in Figure 3. This choice would cause
the node to behave as a CCC node, with the attendant stable load transfer properties
and absence of anchorage problems. A void was also left between the underside of
the column and the top of the test floor to ensure that all of the applied vertical load
was resisted by the connection between the column and footing, because none could
pass in bearing to the platen of the test machine under the footing.

䋭341䋭

FIGURE 4. SPECMEN SF-1: SECTION. (1 inch = 25 mm)

FIGURE 5. SPECIMEN SF-1: PLAN VIEW. (1 inch = 25 mm).

䋭342䋭Figure 5 also shows sets of diagonal bars in the footing. These were placed
both to provide some reinforcement in the otherwise unreinforced corners of thesquare region of cast-in-place concrete surrounding the octagonal column, and toprovide a tension capacity across the pc-cip interface for the purpose of generating
shear friction resistance there. Last, the footing of Specimen SF-1 contained the fullcomplement of transverse reinforcement required by the Caltrans Criteria (2006).
That reinforcement is intended to resist joint shear forces. It should be noted that,
unlike a beam-column joint in a building frame, the region that constitutes the joint,
and in which the reinforcement can be placed, includes a region of the footing outside
the column itself.
Specimen SF-2 was similar to SF-1 except that no footing steel was placed
under the column, thereby eliminating the slots in the bottom of the column. The
same total amount of footing steel was used, but some bars were moved to justoutside the column where they were bundled with other bars already in that location.
The diagonal shear friction steel was also reduced to a single set of bars, rather than
three sets, and the amount of transverse steel in the footing was halved. The basis for
reducing the diagonal steel was that, in both specimens, the normal force due to theflexural steel was ignored in evaluating the shear friction resistance, even though it
appears logical to count it. (This is evident from the strut and tie model in Figure 3b).
The transverse steel was reduced because the tests underlying the Caltrans
Requirements were all conducted on cast-in-place systems in which the column bars
were bent out. That arrangement prevents formation of the strut and tie model of
Figure 3b, in which case the forces must follow a more complex path, such as that in
Figure 3a (from Xiao et al. 1996). It was hypothesized that replacement of the bentout bars by anchor heads in the present study would reduce or eliminate the need for
additional transverse reinforcement.
Specimen SF-3 was designed and constructed after testing SF-1 and SF-2.
Because those two suffered essentially no damage in the connection region, they
provided only lower bounds on the connection strength. To obtain an upper bound as
well (and therefore to bracket the true value), failure in Specimen SF-3 had to beforced into the connection region. To do that, it was designed with a 20-in. (500 mm)
diameter column, as in Specimens SF-1 and SF-2, but a footing that was only 10 in.
(250 mm) thick. The column steel remained the same, but the flexural steel in thefooting had to be much heavier to provide the same flexural strength with a smaller
lever arm.
To avoid a spurious one-way (“beam”) shear failure, a small number offooting ties were needed for one-way shear strength. They were placed so that they
contributed to one-way shear resistance but not to punching shear resistance. A single
set of diagonal “shear friction” steel bars was used, as was done in Specimen SF-2, toprovide a minimum of reinforcement in the otherwise reinforced corners of the region
around the column. The results of tests on Specimens SF-1 and SF-2 had shown thatthe stress in the diagonal bars never exceeded about 5% of yield, so they were notexpected to contribute significantly to shear friction resistance in Specimen SF-3.

䋭343䋭

Test Results
Each test specimen was first subjected to a pure axial load test to investigate
the possibility of shear failure at the precast-cast-in-place interface. The axial loadfor Specimens SF-1 and SF-2 was 240 kips (1077 kN), which was the factored DL +LL on the prototype bridge to be built over I-5, scaled to specimen size. Specimen
SF-3 was subjected to 1.4 times this load. No signs of cracking or damage were seenin any of the three specimens under this loading.
Each test specimen was then subjected to the lateral displacement history
shown in Figure 6. Displacements were applied under stroke control.

FIGURE 6. LATERAL DISPLACEMENT HISTORY.

4000

4000

3000

3000

3000

2000

2000

2000

1000

1000

1000

0
1000

0
1000

2000

2000

3000

3000

4000
15

10

5

0
D ift [%]

5

10

15

Moment [k p- n.]

4000

Moment [k p- n.]Moment [k p- n.]

The load-displacement plots for all three specimens are shown in Figure 7.

4000
15

0
1000
2000
3000

10

5

0
Drift [%]

5

10

15

4000
15

10

5

0
Dri t [%]

5

10

15

FIGURE 7. LOAD-DISPLACEMENT RESPONSE FOR SPECIMENS SF-1,
SF-2 AND SF-3. (1 inch-kip = 113 N-m).
The responses of Specimens SF-1 and SF-2 were nearly identical, andfurthermore they were essentially the same as that of a cip reference specimen tested
earlier (Pang et al. 2010). All the damage occurred in the column, which failed by
combined axial load and flexure in a conventional plastic hinge. The proximate causeof failure was fracture of some of the bars, which was in turn caused by buckling andre-straightening of those bars under cyclic loading. Bar buckling was first observed
at approximately 6% drift ratio in both cases, after which the lateral strength started todrop. The only cracks in the footing were hairline, and all the footing steel displayed
stresses well below yield. The footing was thus behaving as if it were made frommass concrete with no reinforcement.
Specimen SF-3, with the thinner footing, behaved differently. In the earlystages of loading it appeared to be behaving in the same way as the other two, with䋭344䋭

some spalling of the column concrete and yielding of the longitudinal column steel.
However, at about 6% drift, some spalling of the footing concrete became visiblearound the column. The column bars did not buckle or fracture, but the footing
started to sustain more damage. Eventually, at 10% drift, the specimen failed by
combined vertical force and moment transfer in the connection region. Because avoid had been deliberately left under the column to prevent vertical support from thetest floor, the column sank approximately 3 in. (75 mm) through the footing when
failure occurred. Thus the objective of forcing failure to occur in the footing wasachieved. However, the extensive yielding of the column bars and the fact that nofooting damage was visible until quite late in the loading history suggest that the
specimen was only just connection-critical, in which case the ratio of column
diameter to footing thickness used in the specimen represents a fairly tight upperbound on the value needed to avoid footing failure.
Figure 8 shows Specimen SF-3, seen from below, after failure. The column
can be seen to be relatively intact, with all the damage concentrated in thesurrounding footing concrete. The failure surface suggests a punching shear failure
under combined vertical load and bending.

FIGURE 8. SPECEIMEN SF-3 AFTER FAILURE (FROM BELOW).
After the lateral-load testing, Specimens SF-1 and SF-2 were subjected to avertical load test to failure, in order to evaluate the remaining strength of theconnection under vertical load alone. The test could not be applied to Specimen SF-3
because the connection region in it had already failed. In both cases, the column wasable to carry approximately 840 kips (3740 kN) before crushing in the plastic hinge
region of the column. It should be noted that the spiral and several longitudinal bars
had already fractured in the lateral load testing, before this vertical load was applied.
No damage occurred in the footings during this loading. The peak load of 840 kips
(3740 kN) was limited by the axial strength of the previously-damaged column, so thefooting capacity may have been much higher. The load represents 3.5 times thefactored dead plus live load in the prototype bridge, adjusted to laboratory scale. It isthus clear that for footings of these proportions, the shear friction capacity across the
pc-cip interface is easily sufficient to resist the vertical load.

䋭345䋭

Conclusions
The following conclusions were drawn from the study:•

Connection concept. The column-to-footing socket connection can be
designed so that the system behaves like a comparable cast-in-place columnto-footing connection. The precast columns can be designed following thesame specifications as are used to design conventional cast-in-place columns.

•

Need for mechanical anchors. The use in the column of headed straight
bars, instead of bent out bars, simplifies construction and improves the force
flow in the footing, but necessitates the use of headed anchors on the bars.

•

Design against footing failure. The procedures outlined in the AASHTO
Guide Specifications for LRFD Seismic Design for determining the required
flexural strength of the footing were effective in preventing footing failure in
all three column tests.

•

Vertical ties in the footing. When the column steel consists of straight bars
equipped with headed anchors, rather than the conventional bent-out bars, theprescriptive vertical footing ties specified by the AASHTO Guide
Specification perform no useful function and can be omitted. This conclusion
applies only to the prescriptive ties, and not to ties that are needed to supply
shear resistance required to resist computed shear demands.

•

Shear-friction push-through resistance of connection. The strength of theconnection in shear friction was sufficient to prevent any sign of slip, muchless sliding failure, across the pc-cip interface, in any of the three test
specimens. The flexural reinforcement provides normal forces across thepotential sliding interface, which induce sufficient friction to resist thedemands. Thus additional reinforcement (here placed diagonally) is notnecessary for that purpose. However, a small amount of diagonal “trimming”
reinforcement is still desirable to avoid the existence of a large region of
unreinforced concrete at the corners of the embedded column.

Acknowledgements
Support for this work was provided by the Federal Highway Administration, throughtheir Highways for Life program, by the Washington State Department ofTransportation, and by the Valle Foundation at the University of Washington. The
findings and conclusions contained herein are those of the authors alone. Thanks areextended to former graduate students Kyle Steuck and Laila Cohagen, Laboratory
Manager Vince Chaijaroen, Mr. Greg Ritke and the staff of Tri-state Construction,
Mr. Steve Seguirant of Concrete Technology Corporation, Dr. Bijan Khaleghi and
Mr. Jugesh Kapur of the WSDOT, and Dr. Lee Marsh of Berger/ABAM Engineers
for their valuable assistance, without which the project would not have been possible.

䋭346䋭

References“AASHTO (2009). “LRFD Bridge Design Specifications” 4th ed., American
Association of State Highway and Transportation Officials, Washington, DC.
“AASHTO Guide Specification for LRFD Seismic Bridge Design” (2009). AASHTO,
Washington DC.
Building Seismic Safety Council for the FEMA. (2004) “NEHRP Recommended
Provisions for Seismic Regulations and for New Buildings and Other Structures(FEMA 450) 2003 Ed.,” Washington D.C.
Caltrans (2006). “Seismic Design Criteria Version 1.4”. Caltrans, Sacramento, CA.
Matsumoto, E., Waggoner, M., Sumen, G., Kreger, M., Wood, S., and Breen, J.
(2001). “Development of a Precast Bent Cap System,” Center for Transportation
Research, Research Project 0-1748, University of Texas at Austin.
Pang, J. B.K., Eberhard, M.O. and Stanton J.F. (2010). “Large-Bar Connection for
Precast Bridge Bents in Seismic Regions”. ASCE, Jo Bridge Eng. 15 (3) May-Jun:
231-239.
Steuck, K., Stanton, J.F. and Eberhard, M.O. (2009) "Anchorage of Large-Diameter
Reinforcing Bars in Ducts," ACI Structural Journal, July-August, pp 506-513.
Wacker, J., Hieber, D., Stanton, J.F., and Eberhard, M.O. (2005). “Design of Precast
Concrete Piers for Rapid Bridge Construction in Seismic Regions,” Washington
State Transportation Center Report, Seattle, WA.
Xiao, Y., Priestley, M.J.N., and Seible, F. (1996). “Seismic Assessment and Retrofitof Bridge Column Footings”. ACI Structural Journal, ACI, Farmington Hills, MI.
Jan-Feb., pp. 79-94.

䋭347䋭

䋭348䋭

RESEARCH AND REPAIR OF REINFORCEMENT BY OVERLAYINGFLOOR SLAB METHOD IN EXPRESSWAY+LGHND]X +D\DVKLAbstract
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䋭382䋭

27th US-Japan
Bridge Engineering Workshop

Appendix 1:
Japan Participants

Page Left Blank

䋭384䋭

Taku Hanai
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: hanai@pwri.go.jp

Hidekazu Hayashi
Researcher
Road Research Department Bridge Division
Nippon Expressway Research Institute Company Limited
1-4-1, Tadao, Machida-shi, Tokyo, 194-8508, Japan
Phone: +81-42-791-1625
Fax:+81-42-791-2380
E-mail: h.hayashi.ac@ri-nexco.co.jp

Kunihiro Hayashi
Chief Engineer of Osaka Business and Maintenance Department,
Hanshin Expressway Company Limited, Osaka
3-1-25 Ishida, Minato-ku, Osaka, 552-0006 JAPAN
Phone: +81-6-6576-3881
Fax:+81-6-6576-1907
Email: kunihiro-hayashi@hanshin-exp.co.jp

Jun-ichi Hoshikuma
Chief Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6793
Fax:+81-29-879-6739
E-mail: hosikuma@pwri.go.jp

䋭385䋭

Masahiro Ishida
Chief Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6793
Fax:+81-29-879-6739
E-mail: ishida@pwri.go.jp

Shojiro Kataoka
Senior Researcher
Earthquake Disaster Prevention Division
National Institute for Land and Infrastructure Management (NILIM)
Ministry of Land, Infrastructure, Transport and Tourism (MLIT)
1 Asahi, Tsukuba 305-0804, Japan
Phone: +81-29-864-7682
Fax:+81-29-864-0598
Email: kataoka-s92rc@nilim.go.jp

Kazuhiko Kawashima.
Professor
Department of Civil Engineering
Tokyo Institute of Technology
M1-10, O-Okayama,Meguro, Tokyo, 152-8552, Japan
Phone: +81-3-5734-2922
Fax:+81-3-5734-3810
E-mail: kawashima.k.ae@m.titech.ac.jp

Yoshitomi Kimura
Chief Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6793
Fax:+81-29-879-6739
E-mail: y-kimura@pwri.go.jp

䋭386䋭

Kenji Kosa
Professor of Civil Engineering
Kyushu Institute of Technology
Department of Civil Engineering
Kyushu Institute of Technology
Sensui 1-1, Tobata, Kitakyushu, 804-8550 JAPAN
Phone & Fax: +81-93-884-3123
E-mail: kosa@civil.kyutech.ac.jp

Tetsurou Kuwabara
Director of Bridge and Structural Engineering Research Group,Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6 Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516 Japan
Phone: +81-29-879-6793
Fax:+81-29-879-6739
E-mail t-kuwa@pwri.go.jp

Koji Mitani
President and CEO, NEXCO-West USA, Inc.
1015, 18th Street, N.W., Suite 504
Washington, D.C., 20036
Phone: 202- 223-7040
Fax: : 202- 296-5373
Email: k.mitani@w-nexco-usa.com

Jun Murakoshi
Chief Researcher,
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail:murakosi@pwri.go.jp

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Shunji Nagahashi
Staff of Bridges Department, Public Works Division,
Public Works Bureau of Osaka City GovernmentATC-building ITM 6F,2-1-10 Nankokita Suminoe-Ku Osaka City 559-0034,Japan
Phone: +81-6- 6615-6824
Fax:+81-6- 6615-6582
E-mail: s-nagahashi@city.osaka.lg.jp

Keita Nakasu
Senior Researcher,
Bridge and Structures Division, Road Department,
National Institute for Land and Infrastructure Management (NILIM)
Ministry of Land, Infrastructure and Transport (MLIT)
1 Asahi, Tsukuba, Ibaraki, 305-0804 JAPAN
Phone: +81-29-864-4919
Fax:+81-29-864-0178
Email: nakasu-k92gy@nilim.go.jp

Haruhiko Nakata
Inspection and Survey Division, Engineering Department,
Hanshin Expressway Engineering Company Limited
ORIX building 12F, 1-4-1, Nishi-hommachi, Nishi-ku, Osaka City
Phone: +81-6- 6110-7200
Fax:+81-6- 6110-7300
Email: haruhiko-nakata@hex-eng.co.jp

Shoichi Nakatani
Chief Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: nakatani@pwri.go.jp

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Toshiaki Nanazawa
Principal Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
Email: nanazawa@pwri.go.jp

Hideaki Nishida
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
Email: nisida@pwri.go.jp

Fumihiko Nomura
Researcher,
Bridge and Structures Division, Road Department,
National Institute for Land and Infrastructure Management,
Ministry of Land, Infrastructure and Transport (MLIT)
1 Asahi, Tsukuba, Ibaraki, 305-0804 JAPAN
Phone: +81-29-864-4919
Fax:+81-29-864-0178
Email: nomura-f92td@nilim.go.jp

Kiyoshi Ono
Associate Professor of Civil Engineering
Osaka University, Department of Civil Engineering
Graduate school of Engineering
Osaka University
2-1 Yamadaoka, Suita City, Osaka 565-0871, Japan
Phone: +81-6-6879-7598
Fax:+81-6-6871-7601
E-mail: k-ono@civil.eng.osaka-u.ac.jp

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Junichi Sakai
Senior Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: sakai55@pwri.go.jp

Shuhei Sakai
Central Nippon Expressway Co., Ltd.
Dept. of Planning and Design Team Nagoya B.O. NEXCO Central
Mitsui-Sumitomo Bank Nagoya Building
2-18-19 Nishiki Naka-ku Nagoyashi Aichi Prefecture 460-0003, JAPAN
Phone: +81- 52-222-1594
Fax:+81- 52-232-3718
Email: s.sakai.ac@c-nexco.co.jp

Hiroshi Sato
AdvisorIHI Infrastructure Systems Co., Ltd.
Ohama-Nishimachi 3, Sakai-ku, Sakai-shi, Osaka, 590-0977, Japan
Phone: +81-72-223-0980
Fax:+81-72-223-0854
E-mail: hiroshi_sato@iis.ihi.co.jp

Hiroki Sugiyama
Chief
Design Management and Engineering Research Group,Engineering Department
Hanshin Expressway Company Limited, Osaka, Japan
4-1-3, Kyutaro-machi, Chuo-Ku, Osaka, 541-0056, Japan
Phone: +81-6-4963-5790
Fax:+81-6-6252-4583
Email: hiroki-sugiyiama@hanshin-exp.co.jp

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Takashi Tamakoshi
Head,
Bridge and Structures Division, Road Department
National Institute for Land and Infrastructure Management
Ministry of Land, Infrastructure, Transport and Tourism (MLIT)
1 Asahi, Tsukuba, Ibaraki, 305-0804 JAPAN
Phone: +81-29-864-4919
Fax:+81-29-864-0178
Email: tamakoshi-t92gd@nilim.go.jp

Keiichi Tamura
Research Coordinator for Earthquake Engineering
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6 Minamihara, Tsukuba-shi, Ibaraki-ken 305-8516 Japan
Phone: +81-29-879-6721,
Fax:+81-29-879-6739
E-mail: ke-tamura@pwri.go.jp

Yoshiki Tanaka
Senior Research Engineer
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: tanakay@pwri.go.jp

Yukihiro Tsukada
Director of Road Department,
National Institute for Land and Infrastructure Management,
Ministry of Land, Infrastructure, Transport and Tourism (MLIT)
1 Asahi, Tsukuba, Ibaraki, 305-0804 JAPAN
Phone: +81-29-864-4919
Fax:+81-29-864-0178
Email: tsukada-y27x@nilim.go.jp

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Shigeki Unjoh
Research Coordinator for Earthquake Disaster Prevention
National Institute for Land and Infrastructure Management (NILIM)
Asahi 1, Tsukuba-shi, Ibaraki-ken 305-0804 Japan
Phone: +81-29-864-4963
E-mail unjoh@pwri.go.jp

Eiki Yamaguchi
Professor of Civil Engineering
Dept. of Civil Engineering
Kyushu Institute of Technology
Kitakyushu, 804-8550, Japan
Phone: +81-93-884-3110
Fax:+81-93-884-3100
E-mail: yamaguch@civil.kyutech.ac.jp

Guangfeng Zhang
Researcher
Bridge and Structural Technology Research Group
Center for Advanced Engineering Structural Assessment and Research
Public Works Research Institute (PWRI)
1-6, Minamihara, Tsukuba, Ibaraki, 305-8516, Japan
Phone: +81-29-879-6773
Fax:+81-29-879-6739
E-mail: tyou44@pwri.go.jp

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27th US-Japan
Bridge Engineering Workshop

Appendix 2:
US Participants

Page Left Blank

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Ian Buckle
Professor
Department of Civil and Environmental Engineering
University of Nevada, Reno
1664 N. Virginia St/MS 258
Reno, NV 89557-0258
Phone: 775-784-4213,
Fax:
775-784-1390
E-mail: igbuckle@unr.edu

Genda Chen
Professor of Civil Engineering and Associate Director
Mid-America Transportation Center
Department of Civil, Architectural, and Environmental Engineering
Missouri University of Science and Technology
328 Butler-Carlton Hall 1401 N. Pine Street
Rolla, MO 65409-0030
Phone: 573-341-4462
Fax:
573-341-6215
Email: gchen@mst.edu

Waseem Dekelbab
Senior Program Officer
National Cooperative Highway Research Program
500 Fifth Street, NW
Washington, DC 20001
Phone: 202-334-1409
Fax: 202-334-2006
Email: WDekelbab@nas.edu

Ian M. Friedland
Acting Director, Office of Freight Operations and Management
Federal Highway Administration
Phone: 202-366-9615
Email: ian.friedland@dot.gov

John L. Gross
Director of Structures Research, NFRLNational Institute of Standards and Technology
100 Bureau Drive, Stop 8611
Gaithersburg, MD 20899-8611
Phone: 301-975-6068
Email: john.gross@nist.gov

䋭395䋭

Bruce Johnson
State Bridge Engineer
Oregon DOT Bridge Section
4040 Fairview Industrial Dr SE, MS#$
Salem, OR 97302
Phone: 503-986-3344
Email: Bruce.V.Johnson@odot.state.or.us

Kevin R. Mackie
Assistant Professor
Civil, Environmental, and Construction Engineering
University of Central Florida
Orlando, FL 32816-2450
Phone: 407-823-2857
Fax: 407 823-3315
Email: Kevin.Mackie@ucf.edu

Lee Marsh
Senior Project Manager
BergerABAM
33301 Ninth Avenue South, Suite 300
Federal Way, Washington 98003-2600
Phone: 206-431-2340
Email: lee.marsh@abam.com

David Sanders
Professor
Department of Civil and Environmental Engineering
University of Nevada, Reno
1664 N. Virginia St/MS 258
Reno, NV 89557-0258
Phone: 775-784-4288
Fax:
775-784-1390
Email: sanders@unr.edu

Sri SritharanWilson Engineering Professor
376 Town Engineering Bldg.
Civil, Construction & Environmental Engineering
Iowa State University
Ames, IA 50011-3232
Phone: 515-294-523
Fax:
515-294-8216
Email: sri@iastate.edu

䋭396䋭John Stanton
Professor of Civil Engineering
214 More Hall, Box 352700
University of Washington,
Seattle, WA 98195.
Phone: 206-543-6057
Email: stanton@u.washington.edu

Peter Weykamp
Bridge Maintenance Program Engineer
New York State Department of Transportation
50 Wolf Road, Pod 5-1
Albany, NY 12232
Phone: 518-457-8485
Fax:
518-457-4203,
Email:pweykamp@dot.state.ny.us

W. Phillip Yen
Principal Bridge Engineer - Structural Dynamics
Office of Bridge Technology, FHWA
1200 New Jersey Avenue, S.E., Room 73-421
Washington, DC 20590
Phone: 202-366-5604
Email: Wen-huei.yen@dot.gov

䋭397䋭