Proceedings of the International Symposium on Engineering Lessons Learned fromthe 2011 Great East Japan Earthquake, March 1-4, 2012, Tokyo, Japan

22 FEBRUARY 2011 CHRISTCHURCH EARTHQUAKEAND IMPLICATIONS FOR THE DESIGN OF
CONCRETE STRUCTURES
Kenneth J. ELWOOD1, Stefano PAMPANIN2, Weng Yuen KAM3
1

Associate Professor, Department of Civil Engineering, University of British Columbia,
Vancouver, Canada, elwood@civil.ubc.ca
2
Associate Professor (Reader), Department of Civil and Natural Resources Engineering,
University of Canterbury, Christchurch, New Zealand, stefano.pampanin@canterbury.ac.nz
3Structural Engineer, Beca Carter, Hollings and Ferner Ltd, Auckland, kam.wengyuen@beca.com

ABSTRACT: At 12:51pm local time on 22 February 2011, a Mw 6.2 aftershock of theSeptember 4, 2010, Darfield Earthquake shook the city of Christchurch, New Zealand.
The aftershock occurred on an unmapped fault less than 8 km from the city center
resulting in the collapse of two reinforced concrete office buildings and one concrete
parking garage, and severe damage to numerous others. This paper summarizes theobserved damage to concrete buildings in Central Business District (CBD), with a
specific focus on identifying future research to support possible changes to international
seismic design codes to address the Christchurch observations.
Key Words: Christchurch earthquake, ground motions, concrete buildings, collapse,
code changes, shear walls, precast moment frames, repair costs.

INTRODUCTION
Six months after the 4 September 2010 Mw 7.1 Darfield (Canterbury) earthquake, the Mw 6.2
Christchurch earthquake struck Christchurch, New Zealand on the 22 February 2011. The Mw 6.2
earthquake occurred on a previously unknown fault less than 10km south-east of the Christchurch
central business district (CBD), initiating at a shallow depth of 5km. Unlike the 4 Sept event, when
limited-to-moderate damage was observed in engineered reinforced concrete (RC) buildings (Kam etal. 2010), after the 22 February event about 16 % out of 833 RC buildings in the Christchurch CBD
were severely damaged. Whilst there was no fatality in 4 September earthquake (also due to the timeof occurrence i.e. at 4.35am), there were 182 fatalities in the 22 February earthquake (occurring at
12.51pm), 135 of which were the unfortunate consequences of the complete collapse of two mid-rise
RC buildings.
This paper highlights important observations of damage to RC buildings in the 22 February 2011
Christchurch earthquake. Focus will be on damage to modern concrete frame and wall building andthe identification of future research directions necessary to develop appropriate concrete codeprovisions to address the observations in Christchurch. Detailed description of damage to concretebuildings of all vintages and types can be found in Kam et al. (2011).

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GROUND MOTIONS
Recorded ground shaking in the CBD matched or exceeded the 2500 year return period design
spectrum from the NZ Standard (NZS1170.5 2004), particularly in the east-west direction (Figure 1).
Using NZS 1170.5, normal buildings are designed for the 500 year return period design spectrum
using ultimate limit states design principles, while post-disaster buildings are designed using the 2500
year design spectrum shown in Figure 1.
From disaggregation of the seismic hazard, a Mw 6.2 earthquake at a distance of 10 km did notcontribute significantly to the probabilistic seismic hazard model used for Christchurch (Stirling et al2002). With this in mind it raises the question if considering relatively large area sources inprobabilistic seismic hazard assessment to account for earthquakes on unknown faults is sufficient tocapture the risk unknown faults may pose to our urban regions. Should we consider deterministic
scenarios (i.e., shaking from a specific magnitude at a specific distance) for seismically active regions
where unknown faults may dominate the earthquake risk? Research is required on the mostappropriate means of accounting for seismic hazard posed by earthquakes on unknown faults.
One unique aspect of the 22 February earthquake was the very high vertical ground accelerations,frequently exceeding the peak horizontal accelerations at recording station within approximately 10
km of the epicentre. Similar to past earthquakes, the vertical accelerations were characterised byvery high frequency content and peak values were only attained for a very short duration. Research isneeded to define the importance of vertical excitations on building performance considering the highfrequency content and phasing with lateral demands. Rapid attenuation of high frequency verticalground motion should also be considered as this may only be critical for near-fault events.
N
EQ3:REHS(S88E)

1.6

1.4

EQ4:CCCC (N89W)

1.2

Mean of 4
CBD records

1.0

EQ1:CBGS

EQ4:CCCC

EQ2:CHHC

EQ1:CBGS(NS64E)

0.8

EQ2:CHHC (S89W)
NZS1170:5 (2004)
2500−year motion

0.6

NZS4203 (1976 / 1984)
Equivalent elastic

0.4
0.2

NZS1170:5 (2004)
500−year motion

0.0
0

0.5

1

1.5

N

1.8

EQ3:REHS

Principal direction

Spectra Acceleration / S a (g ms-2) .

Spectra Acceleration / S a (g ms-2) .

1.8

1.6
1.4

3.5

4

4.5

5

NZS1170:5 (2004)
2500−year motion
EQ4:CCCC (S01W)

EQ1:CBGS

EQ3:REHS (N02E)

EQ2:CHHC

1.2
1.0

EQ4:CCCC

Mean of 4
CBD records

0.8

EQ2:CHHC (N01W)

0.6

NZS4203 (1976 / 1984)
Equivalent elastic

0.4
0.2

EQ1:CBGS (N26W)
NZS1170:5 (2004) 500−year motion

0.0

2 2.5 3
Period (sec)

EQ3:REHS

Secondary horizontal direction

0

0.5

1

1.5

2 2.5 3
Period (sec)

3.5

4

4.5

5

Fig. 1 Elastic horizontal acceleration response spectra (5%-damped) from Christchurch CBD and the
NZS1170.5 design spectra (red) for Christchurch (NZS soil class D, R=20km)

CONCRETE BUILDING TYPES AND DAMAGE STATISTICS
With a population of approximately 390,000, Christchurch is the second largest city in New Zealandand the economic centre of the South Island. As a relatively centralized city, approximately 25% of
the total employment in the city was located in the Christchurch CBD, leading to a concentration ofbuildings over five storeys in the city centre. In the CBD there were 127 buildings with at least six
stories, with the tallest RC building being 22-storey (86 metres). RC frames and RC walls are the mostcommon multi-storey construction types. Out of 183 buildings with more than 5-storeys, 49% are RC
frame buildings, 22% are RC wall buildings, 7.7% are reinforced concrete masonry and 5.5% are RC
frame with infills. Only 9 steel structures with more than 5-storeys were observed in the CBD.
Buildings constructed prior to the introduction of modern seismic codes in mid-1970s are stillprevalent in the Christchurch CBD. Approximately 45% of the total CBD building stock was built

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prior to the 1970s. Of this, 13.8% or 188 pre-1970s buildings are with 3-storey and more.
Precast concrete floor systems have been used for multi-storey RC buildings in New Zealandsince the mid-1960s. From 1980s to present, the majority of multi-storey RC buildings used precast
concrete floors or concrete composite steel deck systems. Ductile precast concrete frames, designedwith wet connections to emulate cast-in-place construction, were introduced in early the 1980s andsoon became the most popular form of construction for RC frames.
Figure 2 provides statistics for building safety evaluation placarding of concrete frame and wallbuildings in the CBD as of June 12, 2011 (i.e. prior to June 13 aftershocks). While not being a refinedmeasure of damage, coloured placarding, following procedures similar to ATC-20 (NZSEE 2009),
provides a general indicator of the distribution of damage to building types and vintages. In theinterest of brevity only concrete frames and walls are reported here; statics for all building types canbe found in Kam et al. (2011). Statistics for frame buildings indicate a relatively consistent level of
performance for frames constructed prior to 1990, with a slight improvement in performance after
1990. In contrast, for wall buildings the percentage of red placards was markedly higher forpost-1990 buildings compared with buildings constructed prior to 1990. Furthermore, the percentage
of green placards remained reasonably consistent for all vintage of wall buildings. The relatively
high level of damage observed in concrete wall buildings will be discussed with reference to specificobservations in the following sections.
After	  1990	
  (46)

1980s	  (68)

Before	  1980	
  (116)

Unknown	
  (127)
14%

Frames

25%

29%

43%

24%

Before	  1980	
  (27)
15%
41%
44%

56%

27%

59%

30%

32%

Walls

20%

41%

10%
32%

1980s	  (19)

After	  1990	
  (28)

58%

29%

21%

24%

Unknown	
  (17)
47%

50%
29%

Fig. 2 Distribution of placards for concrete frame and wall buildings in CBD as of 12 June 2011,sorted by age of construction. Total number of buildings in each group given in parentheses above pie
chart and approximately reflected in pie chart size. (Data source: Christchurch City Council)
When considering the structural performance implied by the statistics in Figure 2, it is importantto recall the level of ground shaking relative to the earthquake demands assumed in design (Figure 1).
Normal buildings designed to NZS 1170.5 and the concrete design code (NZS3101 2006) are expected
to have a “small margin against collapse” for the 2500 year design spectrum, assumed to be 1.8 times
the 500 year design spectrum (King et al. 2003). Considering the high spectral demands (Figure 1) andthat the two concrete office buildings that collapsed were designed in 1960s and 1980s, it might beconcluded from the point of view of collapse and life safety performance that concrete buildings
designed according to recent building codes performed well or as expected during the Christchurch
earthquake.
While potentially satisfying the objectives of the code, the level of damage observed in many
concrete buildings was severe, leading to a large percentage of buildings currently considered
uneconomical to repair. Potentially more than 50% of buildings in the CBD will be demolished as aresult of damage from the February earthquake (EERI 2011). Fig. 3 shows beam hinging observed ina 22-storey moment frame building. The damage was consistent with the prevalent capacity-design1159

philosophy, protecting the columns from damage and concentrating nonlinear response in the beam
hinges. However, wide residual cracks in the beams have raised concerns regarding low-cycle
fatigue of the reinforcement and repairability of the building. Moving forward, the financial risk anddamage acceptance of ductile RC systems may require further considerations. It is not clear if theperformance of this and other buildings in Christchurch is acceptable to society or if society is willingto pay more for better performance in future earthquakes. However, to enable the selection of different
performance levels in the future, the engineering community should use the impacts of theChristchurch earthquake to promote the further development and implementation of
performance-based seismic design approaches (e.g. ATC-58 2011).

Fig. 3 Beam cracking in 22-story RC moment frame (precast emulating cast-in-place)
EXAMPLES OF DAMAGE TO CONCRETE WALL BUILDINGS
Perhaps some of the most important lessons for modern construction from the Christchurch earthquake
relate to the performance of reinforced concrete wall buildings. Most shear walls in CBD buildings
were tall slender walls where, after the 1982 Concrete Code (NZS 3101, 1982), capacity designconcepts were applied to ensure flexural yielding at the base of the wall limited the shear demands andsufficient horizontal reinforcement was provided to avoid shear failure in the plastic hinge zone. While
this design approach appeared to protect against shear failures in modern wall buildings, unexpected
flexural compression and tension failures in numerous shear walls in Christchurch indicate the need tomodify shear wall design provisions to improve the flexural ductility of slender walls. The following
provides a brief summary of some examples of failure modes observed after the 22 February
earthquake including web buckling, boundary zone and web crushing, and boundary zone steel
fracture.
Wall web buckling
Figure 4 shows the overall buckling of one outstanding leg of a V-shaped (or L-shaped) shear wall in a
7-story building. The width of the buckled web was 300 mm, with an unsupported wall height of
2.66m, resulting in a height-to-thickness (slenderness) ratio of 8.9. The boundary zone extended
approximately 1.2 m into the 4m long web. The boundary steel at the damaged end of the wallconsisted of 16-24mm deformed bars confined by 10 mm smooth bars at 120 mm, with a 180 degree
hook on every other longitudinal bar.

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Photo 3
Photo 2
Photo 1

Fig. 4 Seven-storey 1980s office block with compression failure of the V-shaped RC shear wall.
The wall buckled over a height of approximately 1 m and crushing extended over 3 meters into theweb. Horizontal cracks (approximately 1-1.5 mm width) were visible at the buckled end of the web,while inclined cracks in both directions at approximately 45 degrees were apparent in the middle of
the web over the first story height. Well distributed, primarily horizontal, cracking with widths lessthan 1.5 mm, were observed in the lower half of the first story of the generally undamaged flange. The
damage pattern shown in Error! Reference source not found. suggests that the web may haveinitially experienced flexural tension yielding of the boundary steel, followed by buckling of theunsupported web over the relatively short plastic hinge length. The L-shaped cross-section would haveresulted in a deep compression zone with high compression strains at the damaged end of the web wall.
Stability of the compression zone may have been compromised by a reduction in the web out-of-planebending stiffness due to open flexural tension cracks from previous cycles.
Buckling in a wall with a height to thickness ratio of less than 10, a limit used in several
international codes avoid out-of-plane instability, suggests that a limit on h/t may not be enough toprevent wall buckling. Building on the work by Paulay and Priestley (1993), further research isneeded to determine a relationship between wall buckling, length of plastic hinge and axial stiffness ofthe compression zone after bar yielding.
Boundary bar fracture
Fracture of very light longitudinal reinforcement was also noted in modern high-rise buildings. In
some cases (e.g. Figure 5), wide spacing of transverse reinforcement may have led to bar bucklingprior to bar fracture. The architectural design of the building in Figure 5 included numerous walls,
making it possible to achieve the higher base shear required for a low ductility (nominal or limited
ductile) structural system and thus avoiding the need for full ductile detailing. In some lightly
reinforced shear walls only exhibiting symptoms of flexural cracking, bar fracture of longitudinal bars
was detected only after removal of cover concrete. This has significant implications for both design oflightly reinforced walls but also inspection of similar walls after future earthquakes. Engineers
performing post-earthquake assessments need to be cautious when assessing the extent of damage tolightly reinforced shear walls.
Fracture of boundary reinforcement was also observed in the 200 mm thick wall shown in Figure 6.
This 7-meter long wall (coupled with a 2-meter wall) was the primary E-W lateral force resisting
system for an 8-story plus basement condominium. For the bottom four stories the wall was reinforced
with 12 mm deformed bars at 100 mm in both directions, each face. The boundaries, extending 980
mm from each end, were confined with 6mm smooth hoops at 60 mm, supporting at least every otherlongitudinal bar.
As shown in Photo 1 of Figure 6, fracture of at least four of the 12mm end bars occurred at the top1161

of the ground floor. Core concrete generally remained intact in the confined boundary (except where
fracture of bars occurred); however, crushing of the core extended into the unconfined web for
approximately 3 m from the end of the confined region. The crushing in the web exposed spliced
transverse bars, which could not contain the core concrete once the cover was spalled. The damage inthe web extended diagonally downward from the fractured boundary, suggesting that high shear
stresses may have also contributed to the observed damage.
In terms of future code development, the concrete crushing within the web of the wall in Figure 6
suggests that cross ties may be required outside the confined end zones and splices should be avoided
in transverse bars. Additionally, bar fracture of the lightly reinforced walls shown in Figures 5 and 6
suggests that the minimum reinforcement provisions for boundary zones of shear walls should be
reviewed.

Fig. 5 Buckled and fractured bars in lightly reinforced slender RC shear walls.

Fig. 6 Boundary bar fracture and web crushing in modern 8-story apartment building.
Wall crushing
Figure 7 shows severe damage to a shear wall in the 22-story Hotel Grand Chancellor, designed andconstructed in mid 1980s. Shortening of the ground-floor wall by approximately 800 mm led to avisible lean of the building and restricted access to the potential fall zone around the building.
Significant structural layout irregularities influenced the seismic response of the building; mostnotably the east side of the building was cantilevered over an access lane. Furthermore, the seismic
force resisting system for the lower 14 storeys consisted of shear walls, while perimeter moment
frames were used for the upper stories. (Description below focuses on the performance of the damaged

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wall; further details about the building and the complex damage pattern observed can be found in
Dunning Thornton (2011) or Kam et al. (2012).)
Damage shown in Figure 7 indicates that the wall displaced downward along a diagonal failure
plane through the thickness of the wall. The failure plane, extending the full length of the wall,
appeared to initiate at the top of the lap splice in the web vertical reinforcement. The limited hoops in
the boundary appeared to have opened allowing the boundary longitudinal bars to deform with theshortening of the wall. Crushing of concrete was also noted at the top of the lobby wall, likely toaccommodate the out-of-plane movement of the wall as it slid down the diagonal failure plane.
The wall shown in Figure 7 was likely supporting very high axial loads from several sources. First,
the wall supported a disproportionately high tributary area since it acted as a prop for the cantilevered
bay on the east side of the building. Secondly, the corner column of the upper tower perimeter moment
frame would have imparted high axial loads to the wall due to overturning moments, particularly withany bi-directional movement to the south-east. Thirdly, vertical excitation of the cantilever structure
could have exacerbated the axial load on the wall. Finally, the structural wall would have also
attracted in-plane loads due to N-S earthquake excitation, leading to flexural compression stresses on
one end of the wall. Considering the potential for simultaneous compression from all sources of axial
loads described above, it is expected that the combined axial load and bending in the wall likely
exceeded the concrete compression strain capacity given the limited tie reinforcement provided at thebase of the wall.
Some out-of-plane drift of the wall during the earthquake excitation and the plane of weakness
created at the end of the splice of the web vertical reinforcement at the base of the wall, likely further
contributed to the location of failure observed in Figure 7. Future research is needed on the influenceof the out-of-plane movement of shear walls when combined with high axial loads.

Fig. 7 Crushing and out-of-plane movement of shear wall in 22-story Hotel Grand Chancellor (Middle
and right photos from Dunning and Thornton (2011))

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Observed wall crushing failures, including that shown in Figure 7, suggest that it is best to avoid
compression-controlled walls. It should be recognized, however, that the true axial loads on walls
are not well known, in part due to growth of a wall during shaking and the outrigger effect fromgravity columns. Codified limits on wall axial loads are being considered by the New ZealandDepartment of Building and Housing in response to the damage observed in Christchurch (DBH,2011).
DISPLACEMENT COMPATIBILITY
Similar to past earthquake, the Christchurch earthquake demonstrated the need to carefully consider
displacement compatibility in the design of concrete buildings. The following sections highlight
three specific issues related to displacement compatibility observed in Christchurch; namely, collapseof precast concrete stairs, near-unseating state of precast floors, and severe damage of “gravity system”
not detailed for adequate ductility capacity.

Fig. 8 Collapse of precast stairs
Precast Stairs
Precast stair units collapsed in at least four multi-story buildings, and were severely damaged inseveral other cases, as a result of the 22 February 2011 Christchurch Earthquake. Figure 8 showscollapsed stair units from the 18-storey Forsyth Barr Building (described in detail by Beca (2011)) andthe Hotel Grand Chancellor. Stair damage was particularly prevalent in buildings relying on
moment-frames for lateral load resistance, although stair damage was also noted in some shear wallbuildings. Collapse of stairs not only pose an immediate life safety hazard during the earthquake, but
can result in death or injury after the earthquake as occupant attempt to evacuate the building in dark
conditions without prior knowledge of the stair collapses. Loss of egress routes as a result of the stair
collapses necessitated the evacuation of occupants from windows of high-rise buildings following theChristchurch earthquake.
Stairs are typically designed not to resist building deformations during earthquakes. Movement of
the building relative to the stairs is generally accommodated by seismic gaps and seating provided at
one end of the stair unit. The seismic gap and seating support must be sized for the expected drift
demands during an earthquake. The drift demands from the 22 February earthquake exceeded the
500-year design drift demands required by the New Zealand loading standard (NZS 1170.5, 2004) for
most building periods. Review of stair collapses in the Forsyth Barr building (Beca, 2011) indicatesthat the seismic gap provided was not large enough to avoid closure of the seismic gap anddevelopment of compression forces in the precast stair units. Debris or construction imperfections may

1164

have further decreased the seismic gap, increasing the likelihood of compression in the stair units.
Expected compression loads could have resulted in yielding at the landing and shortening of the stair
units (Beca 2011). Upon reversal of drift demands on the building, the shortened stairs wereparticularly vulnerable to unseating and collapse. Limited seating support would have also increased
the likelihood of progressive collapse once collapse was initiated at one story.
Although damage to stairs was noted in some buildings with shear walls, damage observed in
Christchurch suggests that buildings with moment frames may be particularly vulnerable to stair
collapse. Hinging of beams in moment frames leads to shear distortions in a building frame bay
resulting in lengthening of bay diagonal, and hence, more movement at the stair support.
Furthermore, hinging in beams can result in beam elongation, if not sufficiently restrained by thediaphragm, leading to further displacement demands at the stair support. Stair collapses in the
Clarendon Towers, where precast diaphragms detached from the exterior frames (see Figure 9), havebeen, in part, attributed to the additional deformation demands from beam elongation (Bull 2011).
The poor performance of stairs in the Christchurch earthquake raises several importantconsiderations for the design and evaluation of buildings internationally. Considering the need toprovide egress for occupants of damaged buildings after a major earthquake, it is important that thedrift demand used to size the gap and seating be reflective of that expected in the maximum
considered earthquake. In light of the uncertainty in the ground motions, differences in the linear andnonlinear displacement profile of the building, and the lack of redundancy when seating support is lost,a detail insensitive to construction tolerances or obstructions and allowing for larger than design
displacements should be adopted. Ideally to avoid unintended compression in the stair unit, seismic
gaps should be avoided by allowing the stair unit to slide on the top of the slab surface (Beca 2011).

Fig. 9 Separation of precast floor slabs from supporting perimeter moment frame in 18-storey officebuilding. Photo to right (courtesy of D. Bull) shows temporary supports for columns after earthquake.
Precast diaphragm and beam elongation
Figure 9 illustrates an extreme example in which extensive floor diaphragm damage with near loss of
precast flooring unit supports was accompanied by beam elongation in a perimeter moment frame. The
building shown in Figure 9 is a 17-storey building with ductile perimeter moment frames, internal
gravity frames and flange-hung supported precast double-tee flooring. 60mm topping with cold-drawn
wire mesh reinforcement is used. The perimeter frames have typical 500x850mm deep precast beams
with 600mm square and 800mm square columns. The beam spans are typically 2900mm in the
East-West direction and 5800mm to 6500mm in the North-South direction.
Ductile beam hinging mechanism in the North-South perimeter frames was observed (and repaired)after the 4 September 2010 earthquake. In the 22 February event, the beams in the East-West
perimeter frames experienced hinging. However, as the North-South perimeter frames were previously
hinged and softened, the torsional resistance expected from the overall system would have decreased.
Consequently, the building may have exhibited a moderate level of torsional response, which
amplified the demand on the Northern East-West perimeter frame.

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Due to the high beam depth-to-span ratio (850/2900), the beam elongation effects (geometricalelongation and plastic cyclic cracking) were significantly more pronounced in the East-West perimeter
frames. As expected, the elongation of beams created tension in the connection between the precast
floors and supporting perimeter beams. The largest horizontal crack parallel to the double-tee flange
support was approximately 20mm to 40mm wide. Slab mesh fracture was observed in floor topping
close to the beam plastic hinges. In several locations at the Northern bays, the precast floors have
dropped vertically by approximately 10 to 15mm, indicative of loss of precast floor seating support.
While floor collapse did not occur, the separation of perimeter frames from the diaphragms at
multiple levels in the Clarendon Towers (Figure 9) raises concerns of possible much greater
consequences due to column buckling in aftershocks. To address this risk of collapse in future
aftershocks, pairs of 25 mm rods spanning the full width of the building and attached to spreader
beams were installed at each vulnerable column at four heavily damaged levels (see Figure 9). Where
similar precast floor systems are used, it is important to ensure there is strong connection between thediaphragm and the frames through a well reinforced topping slab in order to ensure separation between
the frames and precast components is minimized. Welded-wire mesh should be avoided in topping forprecast diaphragms given their limited strain capacity. Only ductile welded-wire mesh has been
allowed in New Zealand since 2005 (DBH 2005).

Fig. 10 Damage to gravity columns
Gravity columns
The Christchurch earthquake also reinforced the need to consider deformation compatibility of thegravity system with the seismic force resisting system, particularly for so-called “gravity columns” notassumed to resist lateral loads but essential for gravity load support. Figure 10 shows varying levelsof damage to gravity columns in three different buildings. Photo on right shows one of the criticalweaknesses of the collapsed CTV Building, 400mm diameter columns with 6mm ties at 250mm (forfurther details on the CTV building see Kam et al. (2011)).
The gravity system must be able to accommodate the deformations imposed by the seismic force
resisting system during strong ground shaking. After the axial load failure of several gravity columns
during the 1994 Northridge Earthquake, ACI 318 introduced provisions for “components not part ofthe lateral force resisting system”. For columns deformation compatibility must be explicitly
checked or confinement must be provided to ensure adequate deformation capacity. Similar
provisions have been adopted in other international codes; however, continued research is required
given the challenges of evaluating both deformation demands and capacity accurately.

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CONCLUSIONS
Codes controlling the seismic design of buildings evolve over time, with the greatest advances often
accompanying damage observations from severe earthquakes. Observations from Christchurch are
expected to ultimately impact codes in New Zealand (DBH 2011) and internationally. This paper has
provided examples of observed damage to concrete buildings that relate directly to future researchneeds and potential future changes for seismic design practice and codes.
Some of the important observations and recommendations are summarized below:• The recorded ground motions in the Christchurch CBD are approximately equivalent to the2500 year design spectrum from the NZ Standard, NZS 1170, for many period ranges ofinterest.
• Despite the excessively strong shaking demand, performance of most modern concretebuildings generally exceeded the life safety objectives of the code.
• While many buildings met code objectives, an increasing number of concrete buildings are
now being considered uneconomical to repair, once again raising the question if collapse
prevention is the appropriate target performance level for the building code.
• Research is needed to determine the effects of high frequency vertical ground motion onbuilding structures and how this shaking should be accounted for in design.
• Research is needed to better understand the brittle damage and failure mechanisms observedin walls. Typical wall damage indicates confinement may be required in regions of distributed
web reinforcement and over a height exceeding the assumed plastic hinge length. Similarly
provisions to limit axial load ratio and slenderness ratio should be evaluated.
• In several occasions bar fracture at critical sections was accompanied by limited flexural
cracking. Research is needed to understand this failure mode and develop relevant codeprovisions.
• Displacement compatibility was shown to be a critical issue in (a) collapse of precast concrete
stairs, (b) close-to-unseating state of precast floors due to beam elongation in the moment
resisting frames, and (c) the severe damage of “gravity columns” not detailed for adequate
ductility capacity. Refinement of code provisions is needed to ensure all systems are able towithstand expected displacement demands.

ACKNOWLEDGMENTS
The authors thank Des Bull, Noel Evans, David Hopkins, Steve McCarthy, Chris Van den Bosch, andGrant Wilkinson among many others for their assistance in Christchurch immediately following theearthquake and continued collaboration. The authors are also indebted to Mary Comerio, William
Holmes, and EERI reconnaissance team for many stimulating discussions regarding the lessons forthis important earthquake. Special thanks are also given to the FRST-NHRP project postgraduate
team at University of Canterbury (Dr Umut Akguzel, Sahin Tasligedik, Patricio Quintana-Gallo, Craig
Muir, Joe Bryce) for their fundamental contribution in the field data collection.
The first author is grateful for the financial support for this reconnaissance research from the
Earthquake Engineering Research Institute (through support from an NSF RAPID award) and theCanadian Seismic Research Network (funded by the Natural Science and Engineering ResearchCouncilof
Canada).
The
NaturalHazards
Research
Platform(NHRP)(http://www.naturalhazards.org.nz) short-term Recovery Project funded the field reconnaissance andresearch by the second and third authors. This support is gratefully acknowledged.

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REFERENCES
ATC-58 (2011), “75% Draft Guidelines for the Seismic Performance Assessment of Buildings:
ATC-58-1” Applied Technology Council, Redwood City, California.
Beca (2011). Investigation into the Collapse of the Forsyth Barr Building Stairs on 22nd February
2011. A report to the Department of Building and Housing (DBH). Beca Carter Hollings & FernerLtd., Auckland, New Zealand.
Bull, D. K. (2011). "Stairs and Access Ramps between Floors in Multi-storey Buildings. A report o theCanterbury Earthquakes Royal Commission.", Holmes Consulting Group, Christchurch, N.Z.
DBH (2005), Beware of limitations: Cold-worked wire mesh, Department of Building and Housing(DBH), Wellington, New Zealand.
DBH (2011). Technical Investigation into the Structural Performance of Buildings in Christchurch,
Department of Building and Housing (DBH), Wellington, New Zealand, September 2011.
Dunning Thornton (2011). Report on the structural performance of the Hotel Grand Chancellor in
the earthquake on 22 February 2011. A report to the Department of Building and Housing (DBH).
Dunning Thornton Consultants Ltd., Wellington, New Zealand.
EERI (2011) “The M 6.3 Christchurch, New Zealand, Earthquake of February 22, 2011,” EERI
Newsletter, April 2011.
Kam, W. Y., Pampanin, S., Dhakal, R. P., Gavin, H., and Roeder, C. W. (2010). "Seismic performanceof reinforced concrete buildings in the September 2010 Darfield (Canterbury) earthquakes." Bull.
of New Zealand Soc. of Earthquake Eng., 43(4), 340-350.
Kam, W.Y., Pampanin, S., and Elwood, K. (2011) “Seismic performance of reinforced concretebuildings in the 22 February Christchurch (Lyttelton) earthquake”, Bull. of New Zealand Soc. ofEarthquake Eng., Vol. 44, no. 4, Dec 2011.
King, A.B., Bull, D.K., McVerry, G.H., and Jury, R.D. (2003) “The Australia/New Zealand
Earthquake Loadings Standard, AS/NZS 1170.4” 2003 Pacific Conference on Earthquake
Engineering, 13-15 February 2003, Christchurch, New Zealand: New Zealand Society for
Earthquake Engineering, Paper Number 138.
NZS1170 (2004). NZS 1170:2004 Structural design actions, Standards New Zealand, Wellington, NZ.
NZS3101 (2006). NZS 3101:2006 Concrete structures standard, Standards New Zealand, Wellington,
NZ.
NZSEE (2009). Building safety evaluation during a state of emergency. Guidelines for
territorial authorities. 2nd ed., New Zealand Soc. for Earthquake Eng. (NZSEE),
Wellington, NZ

Paulay, T. and Priestley, M.J.N. (1993). Stability of Ductile Structural Walls, ACI Structural Journal,
American Concrete Institute, Vol. 90, No. 4.

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