Proceedings of the International Symposium on Engineering Lessons Learned fromthe 2011 Great East Japan Earthquake, March 1-4, 2012, Tokyo, JapanDAMAGE OF BRIDGES DUE TO
THE 2011 GREAT EAST JAPAN EARTHQUAKE
Kazuhiko KAWASHIMA
Professor, Department of Civil Engineering, Tokyo Institute of Technology,
Tokyo, Japan, kawashima.k.ae@m.titech.ac.jp

ABSTRACT: This paper presents damage of bridges during the 2011 Great East Japan
earthquake. Since the bridges in the north Miyagi-ken and south Iwate-ken sufferedextensive damage during the 1978 Miyagi-ken-oki earthquake, damage of bridges during
the 2011 Great East Japan earthquake is evaluated in comparison with the damage due tothe 1978 Miyagi-ken-oki earthquake so that the effect of recent progress of seismic
design can be evaluated. Tsunami-induced damage was extensive for bridges along the
Pacific Coast. Typical feature of tsunami-induced damage is presented based on video
movies.
Key Words: Great East Japan earthquake, seismic damage, bridges, ground motions,tsunami, seismic design code

INTRODUCTION
The Great East Japan earthquake (Mw9.0) occurred at 14:46 (local time) on March 11, 2011 along theJapan Trough in the Pacific Ocean. The fault zone extended 450km and 200km in the north-south andwest-east directions, respectively. Extensive damage occurred in a wide region in the east Japan.
As shown later the seismic design codes of bridges were extensively enhanced since 1990. Amonga number of bridges which suffered damage during the 2011 Great East Japan earthquake, the bridges
in the north Miyagi-ken and south Iwate-ken suffered extensive damage during the 1978
Miyagi-ken-oki earthquake. Thus, the Great East Japan earthquake was a valuable occasion to evaluate
the effectiveness of recent progress of seismic design codes by comparing damage due to twoearthquakes. Damage of bridges is shown here for two categories: bridges which were designed inaccordance with the pre-1990 design codes and the post-1990 design codes. Effect of the seismic
retrofit is also presented.
It was the first time to have extensive damage to bridges by tsunami in recent years. No single
word about tsunami is included in the current design codes for bridges. Of course, extensive damage
occurred in the past, but it was probably regarded as unavoidable natural disasters before the WorldWar II. After the World War II, there were tsunami earthquakes, but extensive damage did not occur totransportation facilities.
This paper presents ground-motion-induced damage and tsunami-induced damage of road bridges
in the north Miyagi-ken and south Iwate-ken (refer to Fig. 1) during the 2011 Great East Japan
earthquake.

82

Taro Town
MIYAKO CITY

Tohoku Express
Akita
Express

Otsuchi Town

Kamaishi City
OfunatoCity
Rikuzen-Takada
City
Ichinoseki
City
Kesen-numa
City
Minami-Sanriku
Town
National
Highway 45

Miyako-Kamaishi Area
Yanoura Bridge
Otsuchi Bridge
Namiita Bridge
Horai Bridge
Miyako Bridge
Rikuzen-Takada Area
Nanakiri Overpath
Kesen Bridge
Aneha Bridge
Shiosai Bridge
Kawahara-gawa Bridge
Furukawa Bridge
Numata Overpass

Kesen Area
Nijyu-ichi-hama Bridge
Sodeo-gawa Bridge
Koizumi Bridge

Minami-Sanriku Area
Oritate bridge
Yokotsu Bridge
Onagawa Town Mizujiri Bridge
Minato Bridge
Shizu Bridge
Utasu Bridge
Ishinomaki Area

Ishinomaki City

Sendai City
Natori City

Jyogawa Bridge
Hiyori Bridge
Utsumi Bridge
Shin-Kitakami Bridge

Fig. 1 Bridges in the north Miyagi-ken and the south Iwate-ken

BRIEF HISTORY OF JAPANESE SEISMIC DESIGN FOR BRIDGESFor evaluating damage of bridges, it is important to know what design codes were used forconstructing bridges which suffered damage. Japanese history of seismic design of bridges since theend of 19 century may be classified into the following four stages.
Stage IThe Stage I corresponds to the days between Meiji Revolution and the end of Second World War II(1886-1945) during which seismic design was not considered or was poorly considered. It was the1923 Great Kanto Earthquake when Japanese first realized that technologies imported from theEuropean countries were insufficient for mitigating damage of structures due to an earthquake.
Obviously the technologies imported from the European countries did not include consideration forthe seismic effects. The Great Kanto earthquake was the starting point for Japanese to develop ourown seismic design practice. At the Stage I, damage always resulted from settlement, overturning andexcessive drift of foundations. Countermeasure for damage and cost saving for the use of expensive
steel rebars led to the practice of constructing large and stiff foundations and piers, and this concept
became the main stream of seismic design in Japan.
Stage IIThe Stage II corresponds to the days of the 1964 Niigata earthquake. Extensive damage of bridges
occurred due to liquefaction. Terminology of "liquefaction" came after the Niigata earthquake leading83

to extensive research for the mechanism of liquefaction. The original concept of "unseating prevention
devices" was proposed by Japanese field engineers. They proposed that if adjacent decks were tied
together by cables or if a deck was connected to a substructure, a total collapse of bridges could be
prevented. This concept was first incorporated in the 1971 Design guidelines for seismic design ofbridges (JRA 1971). Now it is widely adopted not only in Japan but also worldwide. There are
essentially no bridges in Japan which do not have unseating prevention devices. Various unseating
prevention devices are currently used.
Stage III
After the Stages I and II, foundations and piers were strengthened, the effect of soil liquefaction was
included in seismic design, and unseating prevention devices were implemented. As a consequence,damage extended in an unexpected direction though the existing damage was mitigated. The days
between 1978 and 1995 when we had extensive damage to piers and steel bearings corresponds toStage III.
During the 1978 Miyagi-ken-oki earthquake (MJMA7.4), extensive damage concentrated to piersand bearings though damage of foundations due to excessive displacement and liquefaction was
gradually mitigated. In the 1982 Urakawa-oki earthquake, extensive shear failure occurred at Shizunai
Bridge at cut-off of main reinforcements with insufficient development. Obviously damage shiftedfrom the previous weak links to the next weak links. Though the allowable shear strength of concrete
was overestimated and the development of longitudinal bars at cut-off points was insufficient, they did
not lead to shear failure of piers in the Stages I and II since the concrete section was large. However as
population in cities increased, a space limitation under bridges restricted the size of piers in viaducts inurban areas. The same restriction was imposed to river crossing bridges for smoother river flow. Thus,
column and pier section had to be reduced in size such that they became flexible piers. Under such a
condition, overestimation of the concrete shear capacity and insufficient development at cut-off
predominantly contributed to resulting in shear failure in columns and piers.
Steel bearings which accommodate only limited relative displacement between superstructure andsubstructure suffered extensive damage. Side blocks with poor capacity were always sheared off, andanchor bolts were pulled out from concrete bases in past earthquakes. Because an elastic static analysis
based on a 0.15-0.3g elastic static seismic force was used, the seismic force demand for steel bearings
was inadequate. There was an argument that weak bearings were fuse to limit an excessive transfer ofthe inertia force from decks to substructures so that collapse of substructures could be prevented.
However damage of weak bearings resulted in excessive drift of superstructures, and repair of steel
bearings in a wide damaged region required long down time resulting in suspension of transportation.
The shift of damage to piers and bearings was first noticed in the 1978 Miyagi-ken-oki earthquake
and it extensively occurred during the 1995 Kobe earthquake.
Stage IV
The importance of considering the realistic design ground motions and ductility capacity by
preventing shear failure for piers and columns was recognized since the 1978 Miyagi-ken-oki
earthquake. The first seismic design code which included the design requirement for ductility capacity
was issued in 1990 as Part V Seismic design, Design specifications of highway bridges (JRA 1990).
An inelastic static analysis as well as the Type I ground motions as shown in Fig. 2 was incorporated
in the 1990 code, where the Type I ground motion represents the ground motions which are induced byan M8 subduction earthquake. In design, the ground motions which were possibly developed in Tokyo
during the 1923 Great Kanto earthquake is used as the Type I ground motions. As well as the Type II
ground motions which was incorporated in the 1995 Guide specifications (JRA 1995) and 1996 code
(JRA 1996), the introduction of realistically high intensity ground motions and an inelastic static
analysis extensively enhanced the seismic performance of brides designed in accordance with thepost-1990 codes. The days after the 1995 Kobe earthquake corresponds to the Stage IV.
Various research for seismic isolation was on going in parallel with the preparation of 1990 design

84

0

4

1971 Seismic design code1980 Part V
1990 Part V

1995 Kobe EQ
2020

3

1996 Part V

2000

1
2
Natural Period (s)502002 Part V

1980

0
0

100

1964 code for steel bridges

0.5

150

1960

1

200

1940

1.5

250

1923 Kanto EQ
1925 First bridge design code

2300

Ground I
Ground II Type I
Ground III
Ground I
Ground II Type II
Ground III

1920

2.5

Number of pages

Response Accleration S (g)code (TRCNLD 1988, PWRI 1992). The first seismic isolated bridge with use of lead rubber bearings(Miyagawa bridge, Shizuoka-ken) and high damping rubber bearings (Yama-age Bridge, Tochigi-ken)
was constructed in 1991 and 1992, respectively. Implementation of elastomeric bearings, LRB and
HDB was initiated in the early 1990s (Unjoh et al 2010). Those bearings were implemented not only
in seismic isolated bridges but also in multi-span continuous bridges for distribution of the inertia
force to piers. The implementation of elastomeric bearings, LRB and HDR extensively mitigateddamage of steel bearings during the 2011 Great East Japan earthquake.
Furthermore, a new evaluation analysis for the inertia force for multi-span continuous bridges wasincorporated in the 1990 code. Prior to the 1990 code, the lateral force was evaluated by only
multiplying a reaction force and a seismic coefficient by disregarding the overall system response. In
the 1996 Part V Seismic design (JRA 1996), the Type II ground motions, which represents typicalnear-field ground motions induced by an M7 event (ground motions developed during the 1995
earthquake), an evaluation for residual displacement (Kawashima, MacRae, Hoshikuma, Nagaya1998), and an enhancement for lateral force demand for bearings and unseating prevention devices
were incorporated (Kawashima 2000, Kawashima 2006a). Thus, the post-1990 design codes (1990code and 1996 code) contributed to construction of bridges with enhanced seismic performance. As aresult, ground-motion-induced-damage of bridges which were designed in accordance with thepost-1990 codes was minor as will be described later.

Year

Fig. 2 Type I and II design ground motions

Fig. 3 Progress of seismic design in terms of numberof pages related to seismic design

Pre-1990 and post-1990 seismic design codes
Fig. 3 shows the progress of seismic design of bridges in terms of number of pages of designcodes which are relate to seismic design. Of course only an increase of number of pages in codes does
not lead to better seismic design, but it can be realized how the knowledge on seismic design was
accumulated in the past. It should be noted that only a 3-4 page description was provided in the 1964
Design specifications of highway bridges, which was referred to in design of a large number of bridges.
Many bridges which collapsed or suffered extensive damage during the 1995 Kobe earthquake wereconstructed in accordance with the 1964 design code (JRA 1964). A combination of a static elastic
analysis and an allowable stress design approach (seismic coefficient method) was used until 1990
(pre-1990 codes). The static elastic method is still used now but a combination of an inelastic static
analysis and the Type I and II design ground motions is the main stream in the post-1990 codes. It
should be noted that elastic and inelastic dynamic response analyses are conducted on routine basis forbridges with complex structural response after 1995.
Seismic retrofit of existing bridges was conducted for reinforced concrete piers which had cut-offs
of longitudinal bars with insufficient development since the 1980s (Kawashima 2006b). Over 30,00085piers were so far retrofitted since 1995. However there still remain a number of piers which require
retrofitting. Moreover, seismic retrofit of foundations has be conducted only for few bridges.

GROUND MOTIONS AND DESIGN GROUND MOTIONS
A number of strong motion accelerations were recorded in the damaged areas by the National Instituteof Earth Science and Disaster Prevention and Japan Meteorological Agency. Since most of them wererecorded on stiff sites, they cannot be directly compared to the Type I and Type II ground motions. Fig.
4 shows some accelerations recorded along the Pacific Coast. Ground accelerations continued over
300s, and had at least two wave groups reflecting the fault rupture process. The highest peak groundacceleration was 27.0 m/s2 which was recorded at Tsukidate City. However this high acceleration was
resulted from a single pulse with high frequency components. The response acceleration at 1.0s period
Acceleration(m/s2)

30
20
10
0

0

100
Time (s)200

Response Acceleration (m/s2)

Fig. 4 Accelerations recorded along the Pacific Coast by NIED
120
100
80

Records at 15 sites
Averaged Response
Acceleration

60
40
20
0
0

1
2
3
Natural Period (s)4Fig. 5 Response accelerations (  =0.05) at 15 sites at the flat land north of Sendai86was only 5.1m/s2 as shown in Fig. 5. Consequently damage of buildings and bridges was very limited
in Tsukidate. This clearly shows that PGA cannot be a reliable index for seismic design.
At soft soil sites in the north Sendai City such as Osaki City, Tome City, Wakuya City and
Ichinoseki City, ground accelerations having higher response accelerations at 0.5-1.5s period range
were recorded. In particular, the response acceleration was slightly over 16 m/s2 at Osaki as shown in
Fig. 6. Thus it is considered that the response accelerations at high intensity areas during the 2011
Great East Japan earthquake was close to but smaller than the Type II design ground motions.
20

NS

Response Acceleration (m/s2)

Acceleration (m/s2)

5
0
–5
5

EW

0
–5
2
0
–2
0

UD

5
10
Time (s)(a) Accelerations during 10s around the peak values

NS
EW
UD

15
10
5
0
0

1

2
3
Natural Period (s)(b) Response accelerations

4

Fig. 6 Acceleration record at Furukawa, Osaki City (K-NET)DAMAGE DUE TO GROUND MOTIONS
Damage of bridges which were constructed in accordance with the pre-1990 codes
Extensive damage occurred at the bridges which were designed in accordance with the pre-1995
design code and not yet retrofitted in accordance with the post-1990 design codes. For example, Photo
1 shows flexural-shear failure of reinforced concrete piers at Fuji Bridge. The damage was resulted
from an overestimated shear capacity and an inadequate development of longitudinal bars at cut-off
which were the common practice in the pre-1980 design codes. This mode of damage occurred
extensively during the 1995 Kobe, Japan earthquake (Kawashima and Unjoh 1997). Extensive
investigation was directed to clarify the failure mechanism of such damage (for example, Kawashima,
Unjoh and Hoshikuma 1995), including a series of large scale shake table experiments using
E-Defense (Kawashima et al 2009). It should be note that damage progresses very sharply once shear
cracks were initiated under this failure mechanism. Seismic retrofit was initiated in the 1980s(Akimoto et al 1990), and it was accelerated after the 1995 Kobe earthquake. Over 30,000 columns
were so far retrofitted. Consequently, during the 2011 Great East Japan earthquake, damage due to thismechanism did not occur at the bridges which were retrofitted, but damage still continued to occur atthe bridges which were not yet retrofitted.
Yuriage Bridge (refer to Photo 2) suffered extensive damage at reinforced concrete hollow andsolid columns, an end of prestressed concrete girders, and steel pin and roller bearings during the 1978
Miyagi-ken-oki earthquake as shown in Photo 3(a). Since the damaged columns were repaired andretrofitted using reinforced concrete jacketing, they did not suffer damage again. However steel pin
and roller bearings suffered extensive damage again in the similar mode as shown in Photo 3 (b). It is
obvious that steel pin and roller bearings are vulnerable to seismic action, because the stress builds up

87

Photo 1 Flexure-shear failure of columns due to termination of longitudinal bars withinsufficient development (Fuji Bridge) (courtesy of Dr. Hoshikuma, J., PWRI)

Photo 2 Yuriage Bridge(b) 2011 Great East Japan earthquake(a) 1978 Miyagi-ken-oki earthquake

Photo 3 Damage of steel roller and pin bearings Yuriage Bridge

to failure by allowing no relative displacements at pin bearings and relative displacements
accommodated in roller bearings are insufficient to real relative displacement under a strong
excitation.
Furthermore, exactly the same end of a prestressed concrete girder which suffered damage during
the 1978 Miyagi-ken-oki earthquake suffered again as shown in Photo 4. It was a critical zone due toconcentration of seismic force, dead load reaction and PC anchor force.
Tennoh Bridge built in 1959 which suffered extensive damage during the 1978 Miyagi-ken-oki
earthquake suffered extensive damage again during the 2011 Great East Japan earthquake at the same88

(b) 2011 Great East Japan earthquake(a) 1978 Miyagi-ken-oki earthquake

Photo 4 Damage of a PC box girder bridge at the end support, Yuriage Bridge

Photo 5 Rupture and buckling of upper braces, Tennoh Bridge, National Road No. 45(a) rupture of a lower brace(b) Disconnection at the end of a lower brace

Photo 6 Rupture of a lower brace and lost of connection at the end, Tennoh Bridge

members. Photos 5 and 6 show rupture and local buckling of upper and lower braces. An end of alower truss brace with a gusset plate was completely disconnected due to corrosion. Because similar
disconnection was observed at other lower braces, it is likely that a large torsion response of the truss89

bridge due to deterioration of torsional rigidity resulted in extensive rupture and buckling of upper andlower braces. This truss bridge was critical for collapse during the earthquake.
Performance of bridges which were retrofitted
Damage of bridges which were already retrofitted suffered virtually no damage. For example, Sendai
Bridge which is an extremely important bridge in Sendai City suffered extensive damage at reinforced
concrete piers and steel bearings as shown in Photo 7(a) during the 1978 Miyagi-ken-oki earthquake.

(a) 1978 Miyagi-ken-oki earthquake
Photo 7(b) 2011 Great East Japan earthquake

Effect of seismic retrofitting of piers, Chiyoda Bridge, National Road No. 4

Photo 8 Elastomeric bearings which were set for replacement of original steel
bearings suffered no damage, Sendai Bridge(a) Rupture of a pin in a pin bearing(b) Pull-out of anchor bolts due to rocking
response of a lower bearing

Photo 9 Damage of steel bearings during the 1978 Miyagi-ken-oki earthquake, Shin-Iino Bridge,
National Road No. 45

90However this bridge suffered no damage during the 2011 Great East Japan earthquake, because
columns were retrofitted as shown in Photo 7(b) and the original steel bearings were replaced withelastomeric bearings as shown in Photo 8.
Shin-Iino-gawa Bridge suffered extensive damage at steel pin and roller bearings as shown in
Photo 9 during the 1978 Miyagi-ken-oki earthquake. It was retrofitted prior to the 2011 Great East
Japan earthquake: 1) some reinforced concrete piers were retrofitted using steel jacketing, 2) nonlinear
viscous dampers were installed, and 3) steel bearings were replaced with elastomeric beatings asshown in Photo 10. As a result, the bridge suffered no damage during the 2011 Great East Japan
earthquake.

(a) An elastomeric bearing without damage(b) A damper for seismic retrofit

Photo 10 Elastomeric bearings and viscous dampers which were set for seismic retrofit did not suffer
damage during the 2011 Great East Japan earthquake

New bridges constructed in accordance with the post-1990 codes
Bridges which were designed in accordance with the post-1990 codes suffered essentially no damage
during the Great East Japan earthquake. For example, Photo 11 shows Shin-Tenno Bridge constructed
in 2002 suffered no damage. Elastomeric bearings and new cable restrainers which satisfy therequirements of the post-1990 design code were set. This bridge was located only 200m upstream of
Tenno Bridge which suffered extensive damage during the Great East Japan earthquake (refer to
Photos 5 and 6).
Photo 12 shows Higashi-Matsushima Bridge which was constructed in 2007. No damage
occurred in this bridge.

(a) Bridge after Great East Japan Earthquake(b)Elastomeric bearing without damage

Photo 11 Shin-Tennoh Bridge91

(a) General view(b) Deck end supported by three elastomeric bearings
Photo 12 Higashi-Matsushima Bridge

Elastomeric bearings generally performed quite well under the extreme ground motions asshown above. However it should be noticed that elastomeric bearings ruptured in some bridges. For
example, several elastomeric bearings ruptured such that a deck offset by 0.5m in the transverse
direction as shown in Photo13(a) at Sendai-Tobu Viaduct. Rubber layers detached from steel plates in
addition to rupture of rubber layers as shown in Photo 13(b). Though detailing is not yet released,
there may be two possible reasons for the damage. The first is a miss design and fabrication of theelastomeric bearings. The second is an interaction between adjacent decks. Since an expansion joint
constrained relative displacement between adjacent decks in the transverse direction, it is possible thata larger displacement demand of a deck is imposed to an adjacent deck resulting in larger shear
deformation in the damaged bearings (Quan and Kawashima 2009).

(a) Offset of left girder by 0.5m due to ruptureof elastomeric bearings (NEXCO East)(b) Rupture of an elastomeric bearing

Photo 13 Damage of elastomeric bearings at Sendai Tobu Expressway

TSUNAMI INDUCED DAMAGE
Bridges which suffered damage by tsunami
A number of bridges suffered damage by tsunami. Overturning of substructures due to scouring did
not occur in road bridges though it happened in railway bridges. Smaller and shorter span bridges

92which were built in the early days were generally vulnerable to tsunami effect. Bridges which weretaller than tsunami waves did not suffer damage. Very short and short span bridges generally sufferedless damage by tsunami probably because tsunami wave front did directly hit bridges and they were
well constraint by abutments.
Back fills and embankment were eroded and washed away at many bridges. Though repair ofback fills and embankment is easier than repair of bridge structures, protection should be considered in
the future.
Utatsu bridge
Utatsu Bridge built in 1972 at Minami-sanriku Town over Irimae Bay suffered extensive damage bytsunami as shown in Photo 14. It was a 303m long 12 simply supported PC girder bridge consisting of3 types of superstructures with spans ranging from 14.4m to 40.7m as shown in Fig. 7. Diaphragms
were set between PC girders at the ends and mid points. The girders had an inclination as large as 6%
in the transverse direction due to curved alignment of the bridge. For example, the inclination was
4.8%, 3%, 3%, 1.1% (sea side down) at D3, D4, D5, and D6, respectively, then it changed to 2-5%,
9%, 4% and 2% (sea side up) at D8, D9 & D10, D11, and D12, respectively.
The bridge was seismically retrofitted a few years ago. Columns were retrofitted by steel
jacketing and unseating prevention devices were installed. The decks D1, D2, D11 and D12 remained
but Decks D3-D10 were washed away. Decks D3-D7 were simply supported pre-tensioned PC girder
bridges. As a part of the seismic retrofit, cable restrainers were set for tying together between D3-D7,
and three steel stoppers were provided at the end of D3 and D7 for preventing excessive longitudinaldeck movements. Though cable restrainers ruptured between D4 and D5, D3-D4 and D5-D6-D7 werestill tied together after floated as shown in Photo 15. D8, D9 and D10 overturned when they were
floated as shown in Photo 16.

A2
D12

D8
D5 D6 D7

D10
P10

D3

P9

D4
D2
P2

P5
P3 P4

D11 P11

D9

P6 P7

P8

SendaiPhoto 14 Utatsu Bridge (Google)

93

Kesen-numa

Locationwhere video
was taken

8.3m

2.3m

(a)

D1-D2
8.3m

8.3m

0.88m
1.75m

(b) D3-D7(c) D8-D12Fig. 7 PC Deck Sections, Utatsu Bridge (the left and right correspond to the land side and sea side,
respectively, when viewed from Sendai side)

Only P2 suffered flexural failure as shown on Photo 17(a) at the land side. The column was
retrofitted using a steel jacket. The damage was probably caused by ground-motion-induced seismic
force based on an analysis of the moment capacity of retrofitted section. Since new ties in the jacket
were flare welded, the ties were still confining the column though they yielded. D2 did not suffer
damage and three stoppers for D2 on P2 were intact. On the other hand, three stoppers for D3 on P2
suffered damage as shown in Photo 17(b). This indicates that D3 offset when D3 was dragged laterally
due to tsunami.
Photo 18 (a) shows two steel longitudinal stoppers on Abutment 2. The main function of the
stoppers was to prevent excessive deck displacement in the longitudinal direction, but they also
restricted deck displacement in the transverse direction. Photo 18 (b) and (c) shows three stoppers andfour pot bearings on P10. Three longitudinal stoppers did not suffer damage at all. All upper bearings
were detached from the lower bearings and floated away together with D10. Three side stoppers in thesea side were removed due to rupture of four anchor bolts each. The sea most lower bearing slightly
uplifted but other two lower bearings were in their original position without damage. From the factthat three stoppers neither suffered damage nor tilted, it was likely that D10 was uplifted over the
stoppers before floated.

Photo 15

Decks D5-D7

Photo 1694

Overturned Deck D8

(a) P2 from D2 side(b) P2 from D3 side
Photo 17 Flexural failure of P2

Photo 18(d) shows three steel longitudinal stoppers and four seat extenders for D9 on P9, andPhoto 18 (e) shows three stoppers for D8 on P7. The stoppers for D8 and D9 neither tilted nor suffereddamage. This also indicates that D8 and D9 were first uplifted before they were floated. Photo 18 (f)shows four stoppers for D7 on P7. The land side stopper slightly tilted with other three stoppers being
not damaged at all. It indicates that D7 (one of the shortest deck) was uplifted toward the land side at
the sea side but insufficiently at the land side before it was floated. Other shorter span decks (D2-D6)
were simply dragged laterally.
A video was taken from A2 by a local resident. This video shows a whole process of rising
tsunami water level until the bridges was completely covered by tsunami as shown in Photo 19. Since
the bridge failed after it was completely covered by tsunami, it is not known when the decks were
uplifted and floated. Since tsunami flow was not fast at both sides of the bay, this saved D1-D2 andD11-D12 from collapse. Tsunami flow velocity was about 6m/s.
Fig. 8 shows possible failure mechanisms due to tsunami. As mentioned earlier, overturning offoundations due to scouring did not occur. The above mentioned damage indicates that Utatsu Bridge
suffered damage due to the mechanism of (b) and (c).
If decks uplift before floated under tsunami action, it may be effective to install restrainers in
the vertical direction between decks and substructures so that decks can be tied down to substructures(tsunami unseating prevention devices). Since installation of tsunami unseating prevention devices
imposes an additional upward force to substructures, substructures have to be strengthen if they do nothave enough capacity. However because an upward uplifting force is limited, such a strengthening is
not generally required for most bridges unless bridges were not very old.

(a) Longitudinal and transverse stopper at A2

95(b) P10(c) Steel bearings and stoppers on P10(d) Steel stoppers and seat extenders on P9(e) P7 from D8 side(f) P7 from D7 side
Photo 18 Damage of stoppers(a) Tsunami reached the bottom of decks

Photo 19(b) The bridge was completely covered bytsunamiVideo taken by a local resident near A2

96

Tsunami

Scoring(a) scoring
Floated after
Tsunami Force uplifted

Tsunami Force

Buoyancy Force(b) Transverse dragFig. 8(c) Floated after uplifted
Possible damage mechanisms by tsunami

Bridges which survived tsunami
There are a number of bridges which survived tsunami though they were completely covered bytsunami. For example, Yanoura Bridge on National Road No. 45 in Kamaishi City was a 108.6m long
three span simply supported curved steel deck girder bridge as shown in Photo 20. It crossed Koshi
River. Pile foundations were used for two abutments and two piers. Ten 30m long and 1m diametercast in place piles were driven to support the footings. A video was taken from Kamaishi Port Office
which was located at the left bank 140m from the bridge as shown in Photo 21. A Toyota rent-a-car
office and Ozawa Building were located near the bridge at the left and right bank, respectively.
At about 15:00, the first tsunami attacked the bridge as shown in Photo 22. Tsunami reachednearly the top of an entrance gate of Toyota rent-a-car office. Since the tsunami reached mid-height ofthe second story of Ozawa Building, it is evaluated that the maximum covering depth of tsunami
above the bridge surface was about 5m. It is noted from the video that few debris was included intsunami which hit the bridge because this bridge was located at the mouth of Koshi River.
As shown in Photo 20, Yanoura Bridge suffered only minor damage on hand rails.

97Kamaishi Port
Construction
OfficeToyota Rent Lease

Yanoura Bridge

downstream

Kohshi-River

Photo 20 Yanoura Bridge, National Road 45, Kamaishi City
50m

Ozawa Building

Photo 21 Location of Yanoura Bridge,
Kamaishi Port Office, Toyota rent-a-car officeand Ozawa Building(a) Tsunami almost covered Yanoura Bridge(b) Yanoura Bridge completely covered bytsunamiPhoto 22 Tsunami attack to Yanoura Bridge

A preliminary evaluation of uplift and floating of Utatsu Bridge
A preliminary analysis was conducted to evaluate possible uplift of decks of Utatsu Bridge. As shownin Photo 16, since the PC girder decks had lateral diaphragms at both ends and mid spans, they werevulnerable to uplift due to trapped air under the deck (Chen 2007). The uplift force by trapped air Fu
was estimated as

Fu  Vta  ww

98

(1)

where, Vta is trapped air volume per deck and ww is unit weight of tsunami water. It was assumed
that ww is 10.78 kN/m3 by considering sand and mud included in tsunami water. Deck weight Wd
was evaluated from the design document.
Table 1(a) shows a comparison of uplift force by trapped air Fu and deck weight Wd for three
sections shown in Fig. 6. Fu is not larger than Wd , but Fu becomes closer to Wd as the deckheight increases (D1-D2 and D8-D12). Thus it is considered possible that D8-D10 were uplifted
before floated if some uplift force due to tsunami additionally applied to the decks.
Similarly, tsunami drag force vs. lateral resistance of a deck was evaluated. The tsunami drag force
Fdf was assumed to be evaluated from hydrodynamic water pressure based on the Design
specifications of highway bridges (JRA 2002, Kosa et al 2010, Shoji et al 2009) asFdf 1 wcd vw2 Ad
2(2)in which  w (  w = ww / g ), g is the gravity acceleration, cd is dag coefficient, vw is tsunami
velocity, and Ad is the side area of a deck. The drag coefficient cd is assumed as 1.4 based on the
Design specifications of highway bridges (JRA 2002). It should be noted that since Eq. (2) representshydrodynamic force acting to a pier, an accuracy of Eq. (2) for representing hydrodynamic force due
to tsunami for a deck is not verified. The inclination of decks in the transverse direction was notconsidered in analysis.
On the other hand, the lateral resistance of a deck was evaluated from the design seismic lateralforce of a deck in the transverse direction Fbr as

Fbr  khWd(3)

where kh is elastic seismic coefficient used in design, Wd is dead weight of a deck and  is anover-strength factor for steel bearings. Since Utatsu Bridges was designed in accordance with thePre-1990 design code, kh was assumed as 0.25. The over-strength factor  was assumed to be 2.0.
Table 1(b) shows the tsunami drag force Fdf and the lateral resistance Fbr of a deck evaluatedfor three types of deck. Since the drag force Fdf is in proportion to the deck height, Fdf becomescloser to the estimated lateral resistance of decks Fbr at D1-D2 and D8-D12. However Fdf is only
45% of Fbr at D3-D7 which were probably dragged by tsunami based on the field investigation.
More precise evaluation for tsunami effect is required.
Table 1A preliminary evaluation for deck uplift and shear resistance(a) Uplift vs. dead weight
Decks
D1-D2
400
Trapped air volume per deck Vta (m3)4300
Estimated uplift force by trapped air per deck Fu (kN)
5800
Deck weight Wd (kN)

D3-D7
55
580
1600

D8-D12
240
2500
3600

(b) Dragged force vs. lateral resistance
Decks
D1-D2
D3-D7
Deck height and length (m)2.5 x 40.7
1.0 x 14.4
2560
360
Hydrodynamic force per deck Fdf (kN)

D8-D12
1.85 x 29.8
1390

Lateral capacity of bearings per deck Fbr (kN)

99

2890

800

1800

CONCLUSIONS

Ground-motion-induced and tsunami-induced damage of road bridges in the north Miyagi-ken andsouth Iwate-ken during the 2011 Great East Japan earthquake was presented. Based on the findingspresented herein, the following conclusions may be tentatively deduced:
1) Ground-motion-induced damage of bridges which were built in accordance with the post-1990
design code was very limited. Thus enhancing the shear and flexural capacity as well as ductility
capacity of piers and extensive implementation of elastomeric bearings were effective for mitigating
damage during this earthquake. Since the ground motions during the 2011 Great East Japan earthquake
was nearly equal or smaller than the type II ground motions, it is the expected level of seismic
performance. However effectiveness of the measures provided in the post-1990 design codes against
stronger than code specified ground motions has to be verified.
2) On the other hand, bridges which were built in accordance with the pre-1990 code and which were
not yet retrofitted suffered similar damage developed during the 1978 Miyagi-ken-oki earthquake.
Appropriate seismic retrofit is required for those bridges.
3) Tsunami-induced-damage was extensive to bridges along the Pacific Coast. There were bridges
which were uplifted before floated. On the other hands, there were a number of bridges whichsurvived though they were completely covered by tsunami. Tsunami effect has to be studied so that it
can be included in design in the future.

ACKNOWLEDGEMENTS

The author express sincere appreciation for a number of organizations and persons for their kind
support and cooperation for damage investigation. Special appreciation is extended to Professor Kazue
Wakamatsu, Professor Kenji Kosa, Prof. Yoshikazu Takahashi, Prof. Mitsuyoshi Akiyama, Dr.
Tsutomu Nishioka, Prof. Gakuho Watanabe, Dr. Hirohisa Koga and Dr. Hiroshi Matsuzaki who weremembers of the Reconnaissance Damage Investigation Team, Japan Society of Civil Engineers. A joint
investigation was also conducted as a part of UJNR Panel on Wind and Seismic Effect activity. Special
appreciation is extended to Dr. Fillip Yen, FHWA and Professor Buckle, I., University of Nevada,Reno, and members of National Institute of Land and Infrastructure and Public Works ResearchInstitute including Drs. Tamura, K., Unjoh, S., and Hoshikuma, J. The author appreciates Mr. Katsuya
Oikawa, resident at Minami-Sanriku Town for proving a video for showing tsunami damage, andProfessor Genda Chen, Professor Jeremy Bricker for introducing various literatures about tsunami.
Financial support by Japan Science and Technology Agency is acknowledged.

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