PACIFIC EARTHQUAKE ENGINEERINGRESEARCH CENTER

Design and Instrumentation of the 2010 E-Defense
Four-Story Reinforced Concrete and Post-Tensioned
Concrete Buildings

Takuya Nagae, Kenichi Tahara, Taizo Matsumori,
Hitoshi Shiohara, Toshimi Kabeyasawa,
Susumu Kono, Minehiro Nishiyama(Japanese Research Team)
andJohn Wallace, Wassim Ghannoum, Jack Moehle, Richard Sause,
Wesley Keller, Zeynep Tuna
(U.S. Research Team)PEER 2011/104
JUNE 2011

Disclaimer
The opinions, findings, and conclusions or recommendations
expressed in this publication are those of the author(s) anddo not necessarily reflect the views of the study sponsor(s)or the Pacific Earthquake Engineering Research Center.

Design and Instrumentation of the 2010 E-Defense
Four-Story Reinforced Concrete and Post-Tensioned
Concrete Buildings

Takuya Nagae, Kenichi Tahara, Taizo Matsumori,
Hitoshi Shiohara, Toshimi Kabeyasawa, Susumu Kono, Minehiro
Nishiyama(Japanese Research Team)

andJohn Wallace, Wassim Ghannoum, Jack Moehle, Richard Sause,
Wesley Keller, Zeynep Tuna
(U.S. Research Team)PEER Report 2011/104
Pacific Earthquake Engineering Research CenterCollege of Engineering
University of California, Berkeley
June 2011

ii

ABSTRACTThis study reports on a collaborative research on the design, instrumentation, and preliminary
analytical studies of two, full-scale, four-story buildings tested simultaneously on the NIED EDefense shake table in December 2010. The two buildings are similar, with the same height andfloor plan; one building utilized a conventional reinforced concrete (RC) structural system withshear walls and moment frames, whereas the other utilized the same systems constructed with
post-tensioned (PT) members. The buildings were subjected to increasing intensity shaking using
the JMA-Kobe record until a near-collapse state was reached. This report summarizes design
issues and design documents, and provides detailed information on the type and location of
sensors used. Initial analytical studies conducted both in the Japan and U.S. to support the design
strategy and instrumentation of the buildings also are documented. The intent of this report is toprovide a resource document for post-test research and high-impact education and outreachefforts.

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iv

ACKNOWLEDGMENTS

Funding for the test program and for Japanese researchers was provided by the Japanese Ministryof Education, Culture, Sports, Science, and Technology. Modest funding was provided by U.S.
National Science Foundation under award number CMMI-1000268 in support of thiscollaboration between U.S. and Japanese researchers.
This report was motivated by the desire to document the importance of these tests and todisseminate the rationale behind this testing program to the broader earthquake engineeringcommunities in Japan and the U.S., as well as other countries, and to highlight important
objectives. The joint report also documents the extraordinary level of collaboration between
Japanese and U.S. researchers studying the response and performance of reinforced concrete
structures. This collaboration has been so incredibly fruitful that universally the authors desire tocontinue such joint efforts in the future for many years to come.
The authors’ wish to acknowledge all the participants within the Reinforced ConcreteGroup of the various NEES–E-Defense workshops held in recent years in Japan and the U.S.
These meetings and the relationships that have developed between the meeting participants havebeen key in laying the foundation for continued strong research collaboration in the present andthe future.
Any opinions, findings, and conclusions or recommendations expressed in this material
are those of the authors and do not necessarily reflect those of the Japanese Ministry of
Education, Culture, Sports, Science, and Technology, the U.S. National Science Foundation, orother individuals mentioned or who have participated in the workshops and meetings.

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CONTENTS

ABSTRACT .................................................................................................................................. iii
ACKNOWLEDGMENTS ............................................................................................................ v
CONTENTS................................................................................................................................. vii
LIST OF FIGURES ..................................................................................................................... xi
LIST OF TABLES ...................................................................................................................... xv
1

INTRODUCTION................................................................................................................ 1
1.1 BACKGROUND ...........................................................................................................1
1.2 OBJECTIVES AND SCOPE .........................................................................................2
1.3 ORGANIZATION .........................................................................................................2
1.4 BRIEF LITERATURE REVIEW AND OVERALL RESEARCHOBJECTIVES ................................................................................................................3
1.4.2

Overall Objectives ............................................................................................ 3

1.4.2

Test Building Specific Objectives .................................................................... 4
1.4.2.1 Performance-Based Seismic Design and Evaluation ......................... 4
1.4.2.1 High-Performance Building with Bonded RC Frame and Unbonded
Post-tensioned Walls .......................................................................... 5
1.4.2.3 Reinforced Concrete Building - Moment Frame Direction ................ 7
1.4.2.4 Reinforced Concrete and Post-tensioned Buildings - Shear Wall
Directions ........................................................................................... 92

TEST BUILDINGS ............................................................................................................ 13
2.1 BACKGROUND .........................................................................................................13
2.2 REINFORCED CONCRETE BUILDING ..................................................................15
2.2.1

Japanese Standard Law Provisions ................................................................. 18

2.2.2

Assessment of RC Building using ASCE 7-05 and ACI 318-08 .................... 19
2.2.2.1 Shear Wall Direction ........................................................................ 19
2.2.2.2 Frame Direction ............................................................................... 23
2.2.2.3 Collapse Mechanism ........................................................................ 33
vii

2.3 POST-TENSIONED BUILDINGS .............................................................................35
2.3.1

Design of Unbonded Post-tensioned Concrete Walls ..................................... 43
2.3.1.1 Performance-Based Design .............................................................. 43

2.4 CONSTRUCTION .......................................................................................................49
3

TEST PLAN AND INSTRUMENTATION..................................................................... 51
3.1 TEST PLAN.................................................................................................................51
3.2 INSTRUMENTATION ...............................................................................................52
3.2.1

General ............................................................................................................ 52

3.2.2

Types of instrumentation ................................................................................ 52
3.2.2.1 Accelerometers ................................................................................. 52
3.2.2.2 Displacement Transducers ............................................................... 54

3.3 GROUND MOTIONS .................................................................................................59
4

SUMMARY, CONCLUSIONS, AND FUTURE WORK............................................... 65
4.1 SUMMARY .................................................................................................................65
4.2 FUTURE STUDIES.....................................................................................................65

REFERENCES............................................................................................................................ 67
APPENDIX A .............................................................................................................................. 71
A.1 MATERIAL PROPERTIES ........................................................................................71
A.2 MEMBER GEOMETRY AND REINFORCEMENT OF THE RC
SPECIMEN ..................................................................................................................73
A.2 MEMBER GEOMETRY AND REINFORCEMENT OF THE PT
SPECIMEN ..................................................................................................................84
A.3 SETUP AND PLACEMENT OF THE SPECIMENS .................................................91
APPENDIX B .............................................................................................................................. 97
B.1 EQUIVALENT LATERAL LOAD PROCEDURE (ASCE 7-05) .............................97
B.2 CALCULATIONS BASED ON ACI 318-08 PROVISIONS ...................................101
APPENDIX C ............................................................................................................................ 133
C.1 CONSTRUCTION PROCESS ..................................................................................133
viii

APPENDIX D ............................................................................................................................ 139
D.1 INSTRUMENTATION .............................................................................................139
APPENDIX E ............................................................................................................................ 227
E.1 PSEUDO ACCELERATION SPECTRA OF THE GROUND MOTIONS..............227
E.2 PSEUDO VELOCITY SPECTRA OF THE GROUND MOTIONS ........................229
E.3 DISPLACEMENT SPECTRA OF THE GROUND MOTIONS ..............................231ix

x

LIST OF FIGURES

Figure 1.1

Elevation of the longitudinal frame. ....................................................................... 6

Figure 2.1

E-Defense shaking table. ...................................................................................... 14

Figure 2.2

Overview of test set up on the shaking table. ....................................................... 14

Figure 2.3

Plan view of specimens......................................................................................... 15

Figure 2.4

Elevation view of specimens. ............................................................................... 17

Figure 2.5

Reinforcement stress-strain relations. ................................................................... 17

Figure 2.6

Concrete stress-strain relations. ............................................................................ 18

Figure 2.7

Equivalent lateral loads on the shear wall system. ............................................... 21

Figure 2.8

P-M interaction diagram for the wall. ................................................................... 21

Figure 2.9

Interstory drift demands for the wall. ................................................................... 23

Figure 2.10

Tributary area for corner column C1. .................................................................... 25

Figure 2.11

Equivalent lateral loads on the frame system. ....................................................... 25

Figure 2.12

P-M interaction diagram for corner column C1. ................................................... 26

Figure 2.13

P-M interaction diagram for interior column C2. .................................................. 27

Figure 2.14

Column shear strength demands. ........................................................................... 28

Figure 2.15

Beam shear strength demands. .............................................................................. 29

Figure 2.16

Column-to-beam strength ratios. ........................................................................... 29

Figure 2.17

Free body diagrams for (a) interior and (b) exterior beam-column
connection. ............................................................................................................ 30

Figure 2.18

Interstory drift demands for the frame system. ..................................................... 31

Figure 2.19

Locations where special hoop requirements are needed. ...................................... 33

Figure 2.20

Collapse mechanism assessment-influence of column yielding level. ................. 34

Figure 2.21

Controlling collapse mechanism in the frame direction. ...................................... 34

Figure 2.22

Controlling collapse mechanism in the wall direction. ......................................... 35

Figure 2.23

Configuration of the steel...................................................................................... 39

Figure 2.24

Hysteretic behavior of cantilever analyses. .......................................................... 42

Figure 2.25

Strength, hysteresis, energy dissipation, and concrete compressive strain at
2% drift angle. ....................................................................................................... 42xiFigure 2.26

Idealized tri-linear lateral load response curve for UPT concrete walls. .............. 44

Figure 2.27

Comparison of experimental and analytical results for test wall TW5 ................. 44

Figure 3.1

Properties of the instrumentation used in the specimens. ..................................... 53

Figure 3.2

Locations of the accelerometers............................................................................ 54

Figure 3.3

Locations of the wire-type displacement transducers. .......................................... 55

Figure 3.4

Locations of the laser-type displacement transducers. ......................................... 56

Figure 3.5

Vertical LVDT configuration (first floor)............................................................. 56

Figure 3.6

Diagonal LVDT configuration (first floor). .......................................................... 57

Figure 3.7

Instrumentation on the reinfored concrete wall. ................................................... 57

Figure 3.8

Strain gauge locations in horizontal and vertical directions at the first floor(reinforced concrete). ............................................................................................ 58

Figure 3.9

Acceleration spectra for JMA-Kobe ground motion (x-direction). ...................... 60

Figure 3.10

Acceleration spectra for JMA-Kobe ground motion (y-direction). ...................... 60

Figure 3.11

Acceleration spectra for Takatori ground motion (x-direction). ........................... 61

Figure 3.12

Acceleration spectra for Takatori ground motion (y-direction). ........................... 61

Figure 3.13

Displacement spectra for the Kobe ground motion (x-direction). ........................ 62

Figure 3.14

Displacement spectra for the Kobe ground motion (y-direction). ........................ 62

Figure 3.15

Displacement spectra for the Takatori ground motion (x-direction). ................... 63

Figure 3.16

Displacement spectra for the Takatori ground motion (y-direction). ................... 63

Figure A.1

Floor plan of the reinfored concrete specimen. .................................................... 73

Figure A.2

Elevation of the reinfored concrete specimen. ...................................................... 73

Figure A.3

Overview of the reinfored concrete specimen. ..................................................... 74

Figure A.4

Details of reinfored concrete specimen................................................................. 76

Figure A.5

Steel locations at floor 1F. .................................................................................... 77

Figure A.6

Steel locations at floor 2F. .................................................................................... 78

Figure A.8

Steel locations at floor 3F. .................................................................................... 80

Figure A.9

Steel locations at floor 3F. .................................................................................... 81

Figure A.10

Steel locations at floor 4F. .................................................................................... 82

Figure A.11

Steel locations at floor 4F. .................................................................................... 83
xii

Figure A.12

Floor plan of the post-tensioned specimen. .......................................................... 84

Figure A.13

Elevation of the post-tensioned specimen............................................................. 84

Figure A.14

Overview of the post-tensioned specimen. ........................................................... 85

Figure A.15

Details of post-tensioned specimen. ..................................................................... 87

Figure A.16

Details of post-tensioned beam column joint........................................................ 88

Figure A.17

Details of post-tensioned wall base and foundation. ............................................ 89

Figure A.18

Details of post-tensioned wall floor slab interface ............................................... 90

Figure A.19

Set up of the specimens......................................................................................... 91

Figure A.20

Placement of the specimens on the shaking table................................................. 92

Figure A.21

Measuring weight of the specimens. ..................................................................... 94

Figure A.22

Weights of equipment on the buildings at the third level ..................................... 95

Figure A.23

Weights of equipment on the buildings at roof level. ........................................... 96

Figure C.1

Construction of reinforced concrete specimen versus post-tensioned
specimen. ............................................................................................................ 133

Figure C.2

Construction of reinforced concrete specimen. .................................................. 134

Figure C.3

Construction of post-tensioned specimen (column). .......................................... 135

Figure C.4

Construction of post-tensioned specimen (beam and slab)................................. 136

Figure C.5

Construction of post-tensioned specimen (walls). .............................................. 137

Figure C.6

Construction of post-tensioned specimen (walls). .............................................. 138

Figure D.1

Measurements. .................................................................................................... 139

Figure E.1

Acceleration spectra for JMA-Kobe ground motion (x-direction). .................... 227

Figure E.2

Acceleration spectra for JMA-Kobe ground motion (y-direction). .................... 228

Figure E.3

Acceleration spectra for Takatori ground motion (x-direction). ......................... 228

Figure E.4

Acceleration spectra for Takatori ground motion (y-direction). ......................... 229

Figure E.5

Pseudo velocity spectra for JMA-Kobe ground motion (x-direction). ............... 229

Figure E.6

Pseudo velocity spectra for JMA-Kobe ground motion (y-direction) ................ 230

Figure E.7

Pseudo velocity spectra for Takatori ground motion (x-direction) ..................... 230

Figure E.8

Pseudo velocity spectra for Takatori ground motion (y-direction) ..................... 231
xiiiFigure E.9

Displacement spectra for the Kobe ground motion (x-direction). ...................... 231

Figure E.10

Displacement spectra for the Kobe ground motion (y-direction) ....................... 232

Figure E.11

Displacement spectra for the Takatori ground motion (x-direction) .................. 232

Figure E.12

Displacement spectra for the Takatori ground motion (y-direction) .................. 233

xiv

LIST OF TABLES

Table 2.1 Weight of RC specimen. ............................................................................................ 16
Table 2.2 Design material properties. ........................................................................................ 17
Table 2.3 Design material properties of post-tensioned specimen. ............................................ 36
Table 2.4 Reinforcement details for PT building. ...................................................................... 38
Table A.1 List of steel reinforcement .......................................................................................... 75
Table A.2 List of steel reinforcement. ......................................................................................... 86

xv

1

Introduction

1.1

BACKGROUND

In the 1994 Northridge and 1995 Hyogo-ken Nanbu (Kobe) earthquakes, many older
reinforced concrete (RC) buildings suffered severe damage, and some collapsed due to brittle
failure of key structural elements. In general, buildings designed to newer standards—such as
the 1981 amendments to Japanese Building Standard Law Enforcement Orders and the 1976
and later versions of the U.S. Uniform Building Code—performed well. Some newer U.S.
buildings performed poorly due to substandard behavior of diaphragms, particularly in precast
prestressed concrete parking structures and gravity systems. In both Japan and the U.S.,
although building response to strong ground shaking generally satisfied code requirements andperformed adequately in providing life safety, high repair costs as a result of nonlinear
behavior produced large member cracks and residual deformations.
As a result, new design approaches were developed that focused on defining
deformation limits that can be used to assess both collapse safety and the impact of damage on
repair costs and loss of building use (down time). In the U.S., these new approaches are
documented in FEMA-356 report and by reports published by the Pacific Earthquake
Engineering Research (PEER) Center and others. Damage observed from significant
earthquakes often results in an evolution of design practice, as witnessed in the 1994
Northridge earthquake for structural steel buildings and in the 2010 Chile earthquake forreinforced concrete wall buildings. As well, there is continuous pressure to develop structural
systems that allow for longer spans and more flexible floor plans using new materials or new
systems, such as prestressed and post-tensioned (PT) concrete systems. These new systems
often have attributes that are different from commonly used systems, where laboratory testingand experience in earthquakes of both components and systems have been used to assess1expected performance and to verify design approaches. For example, PT systems typicallyhave low hysteretic energy dissipation capacity relative to reinforced concrete (RC) systems;
however, this same attribute tends to limit residual deformations. Therefore, it is important tocontinuously assess the expected performance of buildings constructed using new codes andnew systems via testing of large-scale components and full-scale buildings models subjected
to realistic loading histories expected in both frequent and rare earthquakes.

1.2

OBJECTIVES AND SCOPE

A series of shaking table tests were conducted on essentially full-scale RC and PT buildings
designed using the latest code requirements and design recommendations available in both
Japan and the U.S. To assess performance in both moderate-intensity frequent earthquakes(service-level) and large-intensity very rare earthquakes (collapse-level), the buildings were
subjected to increasing intensity shaking using the JMA-Kobe and Takatori records until anear-collapse state was reached. The tests were designed to produce a wealth of data on
stiffness, strength, and damping over a large range of deformations to assess current codes andrecommendations, and will be used to develop new analysis tools and design
recommendations, and determine if limit states and fragility relations used in current
performance-based approaches to limit repair costs and assess collapse are consistent withmeasured responses and observed performance. The tests also will provide a wealth of data toassess and improve existing analytical tools used to model RC and PT components andsystems, as well as help to identify future research needs.

1.3

ORGANIZATIONThis report is divided into four chapters. The first chapter includes a brief introduction andbackground, followed by a short summary of the overall research objectives from both U.S.
and Japan perspective. Chapter 2 provides an overview of the two test buildings, including asummary of design requirements, construction materials, structural drawings, and specimen
construction. Chapter 3 includes a detailed description of the instrumentation used for eachtest building. Chapter 4 provides a brief summary and conclusions, as well as an overview ofplanned future studies.

2

1.4

BRIEF LITERATURE REVIEW AND OVERALL RESEARCH OBJECTIVES

The lengthy planning process and extensive collaboration between U.S. and Japan researchers
leading up to the December 2010 tests produced test buildings that were designed to provide
vital and important behavior and design information for both the U.S. and Japan. Because
design objectives/requirements and performance expectations are somewhat different between
the U.S. and Japan, a more detailed description of specific research objectives is provided in
the following sections. In Chapter 2, the final building designs are reviewed using ASCE 7-05,
ACI 318-08, and ACI ITG 5.1-07 to provide detailed information on U.S. code provisions anddesign recommendations that were met or not met.
1.4.2 Overall Objectives
When the Japanese Building Standard Law Enforcement Orders was substantially updated in
1981, the guiding principles of the new code were to prevent damage in minor and moderate
earthquakes and to prevent collapse in severe earthquakes. These principles are essentially thesame as those embodied in U.S. codes at the time (e.g., the Uniform Building Code).
However, observations based on the 1994 Northridge earthquake and the 1995 Kobe
earthquake, as well as other moderate to strong earthquakes that have occurred in recent years
near major urban cities in Japan, have revealed that many buildings became nonfunctional andnonoperational due to damage to non-structural systems even if the structural damage waslight to moderate. Based on these experiences, new design approaches have emerged in the
1990s and 2000s that address both structural and non-structural damage over a wider range ofhazard levels. These approaches, which differ from prescriptive codes such as Uniform
Building Code or the International Building Code, are commonly referred to as performancebased approaches, since the objective is to provide a more rigorous assessment of building
performance.
Performance-based design approaches also provide a means to communicate
expectations of building performance to the general public, building owners, and government
agencies. This dialogue is essential, as there is a perception among the general public thatbuildings, both in Japan and the U.S., are “earthquake proof.” This perception is inconsistent
with the stated code objectives of collapse avoidance. The economic losses and societal

3impacts associated with buildings designed with current prescriptive code requirements arelikely to be very significant, potentially impacting the affected region for many years.
Novel approaches have emerged to provide improved performance, for example,
approaches that utilize response modification such as base isolation or using dampers.
Although these approaches may offer excellent performance, in general, initial costs are highand other challenges exist (for base isolation one significant hurdle is accommodating therelative movement between the superstructure and the surrounding foundation, including
utilities). Consequently, only a limited number of buildings are constructed utilizing theseapproaches.
Therefore, it is essential to continue developing performance-based approaches in
conjunction with innovative cost-effective building systems that are capable of better
performance relative to conventional construction. The RC and PT Buildings that aredescribed in Chapter 2 were designed and the test protocol developed to provide vital
information to address both of these issues. In the following three subsections, more detailed
descriptions of test objectives are provided.
1.4.2 Test Building Specific Objectives
1.4.2.1 Performance-Based Seismic Design and Evaluation

Application of performance-based seismic design (PBSD), or performance-based seismic
evaluation (PBSE), e.g., based on the PEER framework, has become fairly common. At aminimum, two hazard levels are considered: one associated with fairly frequent earthquakes
with a return period of 25 or 43 years (a service-level event), and one associated with veryrare earthquakes with a return period of approximately 2500 years (the Maximum Considered
Earthquake, or MCE). A comprehensive PBSE might consider many hazard levels, e.g., ATC58 [ATC 2007] considers 11.
Although relatively complex nonlinear modeling approaches are used to model frame
and wall buildings, there is a lack of field and laboratory data available to assess the reliabilityof these models. With respect to shake table testing, data are mostly available for simple
systems with one or two bays and one or two stories, often for effectively two-dimensional,
moderate-scale structures utilizing a single lateral-force-resisting system (references) and4

without gravity-load-resisting systems/members. The test buildings described Chapter 2 and 3
are essentially full-scale, three-dimensional buildings with different lateral-force resisting
systems in the orthogonal directions. The availability of detailed measured response data
along with observed damage will enable comprehensive system-level studies to assess thefollowing issues: (i) the ability of both simple and complex nonlinear models to capture
important global and local responses, including system interactions, both prior to and after
loss of significant lateral strength; (ii) the capability of existing modeling approaches tocapture loss of axial-load-carry-capacity (collapse); and (iii) the reliability of proposed PBSE
approaches for new buildings (e.g., ATC-58) to predict the degree and distribution of damage
and the related repair costs, as well as the margin against collapse for very rare events (e.g.,
MCE or higher level shaking).
1.4.2.1 High-Performance Building with Bonded RC Frame and Unbonded Posttensioned Walls

One approach that improves a building’s performance is self-centering structural systems thatutilize unbonded prestressed tendons. Initial research, conducted as part of the U.S. National
Science Foundation’s (NSF) PREcast Seismic Structural Systems (PRESSS) program in the
1990s [Shiohara 2001; Zhao and Sritharan 2007; Priestley 1991] demonstrated that such
systems sustained relatively low damage compared to conventional RC systems under similarloading. This system has been implemented in a 39-story building in California [Priestley
1996] and for bridges [Priestley et al. 1999]. The self-centering framing system tested by the
PRESSS program involved relatively complex beam-column connection details. Subsequent
research has been conducted to develop alternative systems/details [Englekirk 2002] and toextend the concept to steel structures [Pampanin et al. 2006] and timber structures [Pampanin
2005].
Primary research on self-centering systems in Japan began in 2000, with tests on
hybrid column-beam joints with unbonded prestressing tendons and mild steel inside members
by Sugata and Nakatsuka [2004], which was similar to the U.S. hybrid column-beam joint
system. Sugata and Nakatsuka also proposed a numerical model [2005] to simulate flag shapehysteresis behavior exhibited by these connections, and Niwa et al. [2005] studied unbonded
PT precast column-beam joint with external damping devices under the beam. Ichioka et al.

5[2009] tested PT precast concrete portal frames with a corrugated steel shear panel placed
between the beam and the foundation beam.
As shown in Figure 1.1, shake table testing has been conducted on reduced-scale(25%), three-story PT frames with bonded and unbonded beams [Maruta and Hamada 2010].
Test results demonstrated that PT precast concrete frames were very ductile, yet only minor
damage was observed for velocities less than 50 kine. However, due to the self-centering
capability, the system displayed low energy dissipation capacity (no damping devices wereused). Self-centering systems have been developed and tested for structural steel systems[Ikenaga et al. 2007; Ichioka et al. 2009]; these systems have not yet been used in practicebecause design procedures have not been established to satisfy the Japanese BuildingStandard. In addition, the initial cost for the self-centering system is higher than conventional
RC systems, and the potential long-term benefits of the system have not been sufficiently
studied to assess if the higher initial cost is justified.

(a)Figure 1.1(b)

Elevation of the longitudinal frame [Ikenaga et al. 2007].

In this study the PT concrete structure is denoted at the “PT Building.” The design ofthe building is based on typical Japanese practice, with grouted PT precast prestressed
concrete structure for beams and columns and unbounded prestressed concrete shear walls to
6

provide energy dissipation. To adequately compare the response of the RC Building and the
PT Building, it was mandatory that the PT Building be designed such that the lateral force
capacity of the PT specimen be close to that of RC specimen (for scientific interest); note that
the Japanese code requires that the PT Building have slightly larger lateral strength than the
RC Building. The PT Building also used high-quality, high-strength concrete. The innovative
energy dissipative device utilized in the PT Building—the unbonded PT shear wall—has been
investigated previously (see discussion above), but they have not been used in practice ineither Japan or the U.S.
1.4.2.3 Reinforced Concrete Building - Moment Frame Direction

The conventional RC building system (RC Building) was designed to satisfy typical seismic
design practice in Japan, with the quantity and arrangement of longitudinal and transverse
reinforcement conforming to the Building Standard Law Enforcement Order and AIJStandard. Typical materials were used to construct the test specimen. Preliminary analyticalresults presented by U.S. researchers at the October 2009 meeting in San Francisco and at the
March 2010 meeting in Tokyo indicated that the design also reasonably represented U.S.
Special Moment Frame (SMF) construction in California. A detailed assessment of the RC
Building relative to U.S. code provisions is presented in Chapter 2.
Reinforced concrete special moment-resisting frames (SMRF) are commonly used inseismic regions, particularly for low- to mid-rise construction. Their behavior during seismic
excitation depends on the behavior of individual members (e.g., columns, beams, joints, andslabs) and the interaction between members. Although numerous component tests have been
performed on RC columns [Berry et al. 2004], beam-column joints and slab system tests that
capture the interaction between these elements are rare [e.g., Ghannoum 2007; Panagiotou
2008]. Even less common are system tests that account for multi-directional dynamic loading
effects. The E-Defense tests will help fill the knowledge gap in this area.
The influence of beam-column joint behavior on performance of the RC Building was
identified as a topic of interest that could be assessed with the test buildings. Because test data
within this range were not well represented in the literature and this range of strength ratios iscommon in Japan, Hiraishi et al. [1988] conducted quasi-static tests on beam-column joints
with column-to-beam strength ratios between 1.0 and 2.0. The test results indicated that the7beam-column joint specimens performed uniformly poor, with significant strength loss andseverely pinched hysteresis behavior due to bar slip, even if the demand on the joint (frombeam yielding) was less than the joint shear strength. Given this information, the RC Building
was designed to have beam-column joints that satisfy the weak-beam strong-column concept,
but with calculated column-to-beam strength ratios near 1.2 for interior joints and 1.6 forexterior joints, respectively. The objective was to assess the behavior of joints in aconventional design at full scale on the E-Defense shake table.
As the structural engineering field moves towards PBSD, it is increasingly importantto accurately model the full nonlinear behavior of SMRFs. Many challenges arise in nonlinear
dynamic simulation due to the complex interactions between members and the variability inmember boundary conditions. Current key challenges in simulating the seismic behavior of
SMRFs are summarized below:(1) Evaluating the “elastic” stiffness of all members: Structural stiffness is crucial for
obtaining the correct seismic demand. Member stiffness is variable during seismic
excitation and largely depends on axial load and level of cracking [Elwood andEberhard 2009]. Element interactions also play a vital role. For example, strain
penetration of longitudinal bars of columns and beams into joints and foundations canaffect the stiffness of a structure by as much as 40% [Sezen and Setzler 2008; Zhao
and Sritharan 2007]. Strain penetration effects in joints are highly dependent on joint
demands and confinement, which can only be obtained from system tests.
(2) Evaluating the strength of each member at which its behavior softens significantly:
In SMRF that strength usually coincides with the yield strength. It is particularly
critical to achieve a model with the correct ratios of member strengths so that correctmechanisms are determined. While member yield strength can be estimated withreasonable accuracy for individual columns and beams, it is quite difficult to assessthat strength in complete structural systems, particularly for monolithic beam/slab
systems and joint construction. Quantifying the contribution of the slab on beam andjoint capacities as well as the effect of strain rate effect under dynamic excitation is anespecially important challenge that requires full system tests.

8

(3) Simulating the post-“yield” response of each member: Dynamic tests that cycle a
structural system to very large deformations are necessary to obtain information about
post-yield behavior. Structural assessment for the collapse prevention performance
objective requires the identification of the deformation at which strength degradation is
initiated and the ensuing degrading behavior. Such degradation can be the result of barbuckling, loss of shear strength, and fracture of transverse reinforcement in SMRF.
Loading history and load sharing between structural elements both affect the initiationand the propagation of damage in elements. If adjacent elements are able toredistribute loads the behavior of the failing elements is significantly altered[Ghannoum 2007; Elwood and Moehle 2008]. Component tests cannot capture such
system effects.
(4) Simulating joint deformations and their progression during seismic excitation: Aswith strain penetration effects, joint deformations can significantly affect the lateral
stiffness of a SMRF. The joint-softening effect is particularly high at large
deformations where joint damage can be substantial. The difficulty in assessing joint
behavior stems from the fact that slabs, beams, and columns affect their behavior
substantially. The beam-to-column strength ratio has particular influence on joint
behavior [Shiohara 2001] as does bi-axial loading.
(5) Assessing bi-axial loading effects on columns: very few column tests are performed
under bi-axial loading and even fewer dynamically. Bi-axial loading affects column
strength as well as strength degradation.
1.4.2.4 Reinforced Concrete and Post-tensioned Buildings - Shear Wall Directions

Common Japanese practice uses columns at wall boundaries that are wider than the wall web(so-called barbell-shape). Over the past twenty years in the U.S., however, it has becomecommon practice to design walls with rectangular cross sections. (Based on test results
available in the literature, the AIJ Standard for “Structural Calculations of Reinforced
Concrete Buildings” was revised in 2010 to show RC walls with rectangular cross section.)
Although the deformation capacity attributed to wall shear failure or wall bending
compression failure can be estimated using the "AIJ Design Guide Lines for Earthquake
9Resistant Reinforced Concrete Buildings Based on Inelastic Displacement Concept," theseprocedures can be applied to walls with rectangular cross sections. Therefore, walls withrectangular cross sections were used in both the RC and PT Buildings to assess wall behavior
at full-scale under dynamic loading. Primary objectives of the tests were to assess the behavior
and performance of shear walls with rectangular cross sections to provide data to assess
common practice in the U.S. and to potentially change practice in Japan, as well as to enable aside-by-side comparison between the conventional RC walls and high-performance PT walls.
Behavior and modeling of shear walls has received increased attention in recent years
because not only do shear wall systems provide substantial lateral strength and stiffness, they
are resilient to complete collapse [Wallace et al. 2008; EERI Newsletter 2010]. Recent testing
conducted within the NEES-Research program includes quasi-static testing at: (i) nees@UIUC
on isolated cantilever walls with rectangular cross sections with and without lap splices by
Lowes and Lehman; (ii) nees@Minnesota on isolated, cantilever walls with both rectangular
and T-shaped cross sections subjected to uniaxial and biaxial loading by French and Sritharan,and (iii) nees@UCLA by Wallace and nees@Buffalo by Whittaker on low-to-moderate aspectratio (one to two), isolated walls with rectangular cross sections. Shake table tests on verylarge scale, eight-story walls with both rectangular and T-shaped cross sections subjected touniaxial loading have been conducted at nees@UCSD (Panagiotos and Restrepo). Tests also
have been conducted on PT walls (Sause and others). Therefore, the full-scale shake table
tests on the RC and PT Buildings will provide a wealth of data, including information on shear
wall systems (walls and frames) subjected to three-dimensional, dynamic loading.
Nonlinear modeling of shear walls has been the subject of much research in the last
five years, with considerable attention has focused on modeling flexure-shear interaction, i.e.,
where yielding in shear is observed for relatively slender, isolated walls, with aspect ratios( Aw = hw lw ) between 2.4 (PCA tests) and 3.0 (e.g., see Massone and Wallace [2004]), eventhough the computed nominal shear strength exceeds the shear demand. The RC Building
tested at E-Defense will provide important results for system level tests of slender walls( Aw = 4.8) coupled by a shallow beam to corner columns at low axial load. The tests will
provide data for a case where flexure-shear interaction is expected to be minor. Quasi-static
tests are currently being conducted to assess flexure-shear interaction for moderate aspect10ratio walls ( Aw = 1.5 to 2.0 ) and quasi-static loading [Tran and Wallace 2010]; future shake
table testing is needed to further address this need.
Slightly different detailing has been provided within the yielding regions (plastic hinge
regions) of the shear walls on the north and south sides of the conventional RC building toinvestigate the role of detailing on damageability, lateral strength degradation, and,potentially, the loss of axial load carrying capacity. Given the likely role of detailing on theobserved damage in the recent Mw 8.8 February 27, 2010, earthquake in Chile, this aspect of
the test is of significant interest.
The impact of modest coupling on lateral story displacements and wall shear forces
has not yet been studied, particularly for dynamic loading of three-dimensional building
systems. The E-Defense tests will provide a wealth of data to assess these issues, as well as
the increase in wall shear with shaking intensity.

11

12

2

Test Buildings

Descriptions of the RC and PT buildings are provided in the following sections. Backgroundinformation is provided on the E-Defense shake table and detailed information on overall
geometry, member dimensions, and longitudinal and transverse reinforcement are presented
for the RC and PT buildings.

2.1

BACKGROUND

The E-Defense shake table, the largest in the world, has plan dimensions of 20 m × 15 m
(Figure 2.1). The table can produce a velocity of 2.0 m/sec and a displacement of 1.0 m in two
horizontal directions, simultaneously, and accommodate specimens weighing up to 1200
metric tons. In this study, two four-story buildings were tested, one RC and one PT. The twobuildings were almost identical in geometry and configuration, and were tested
simultaneously, as shown in Figure 2.2. Each building weighed approximately 5900 kN;
therefore the combined weight of the two buildings was 98% of E-Defense table capacity. The
test buildings utilized different structural systems to resist lateral forces in the longitudinal andtransverse directions. In the longitudinal direction, a two-bay moment frame system was used,
whereas in the transverse direction, structural (shear) walls coupled to corner columns by slabbeams were used at each edge of the buildings (Figure 2.3). Story heights at all levels for bothbuildings were 3 m, for an overall height of 12 m. The plan dimensions of the buildings were
14.4 m in the x- or frame direction and 7.2 m in the y- or wall direction.

13

X
Y

20 m

15 m

Figure 2.1 E-Defense shaking table.

Figure 2.2

Overview of test set up on the shaking table.

142.5 m2@7.2 mShaking
table

x

y

7.2 m

Figure 2.3

2.2

Plan view of specimens.

REINFORCED CONCRETE BUILDING

Plan and elevation views of the structure are shown in Figure 2.3 and Figure 2.4, respectively.
Cross-section dimensions of columns were 500 mm × 500 mm, and walls were 250 mm ×2500 mm; beam cross-sections were 300 mm × 600 mm (width × depth) in the x-direction and
300 mm × 400 mm for interior beams and 300 mm × 300 mm for exterior beams in the ydirection. Additional beams with cross sections of 300 × 400 mm supported the floor slab atintervals of 1.5 m in the y-direction. A 130 mm-thick floor slab was used at floor levels 2
through 4 and at the roof level. Detailed information on member geometry and reinforcement
used is given in Appendix A.2. Information on the building weight and material properties arecontained in Table 2. and Table 2., respectively. Building weight was calculated based on thedesign, i.e. before the non-structural members were placed in the specimens. Floors 2 through4 weighed about 900 kN, whereas the weight of the roof was 1000 kN; the remaining weight
was in the foundation. The weight of the equipment is presented in Appendix A.1.
The design concrete compressive strength was 27 N/mm2, with SD345 D19 and D22bars used for primary longitudinal reinforcement. Information on the longitudinal andtransverse reinforcement used in all members is provided in Table 2. and Figure 2.5. Typicalconcrete stress versus strain relations are given in Figure 2.6. See Appendix A.1 for detailed
information on as-tested material properties.

15

Table 2.1

Weight of RC specimen.

RC

Structural

2.4

t/m3

RFL

4FL

3FL

2FL

Base

Column

5.4

10.8

10.8

10.8

5.4

Girder

16.4

16.4

16.4

16.4

216.2

Wall

4.1

8.1

8.1

8.1

4.1

Slab

44.1

43.7

43.3

42.8

10.6

Beam

8.0

8.0

8.0

8.0

0.0

Parapet

5.3

0.0

0.0

0.0

0.0

Steel

Temp. Girder

0.0

0.0

0.0

0.0

0.3

Sum

[t]

83.3

87.0

86.6

86.2

236.5

Stair

330

360

360

360

0

Measurement

0

3000

1750

1690

1690

Handrail

244

271

271

271

197

on the slab

4633

180

0

0

0

under the slab

495

0

0

0

0

RC Base

6042

346

0

0

0

Ceilingunder the slab

296

0

0

0

0

Sum

[kg]

12040

4157

2381

2321

1887

Total

RFL

4FL

3FL

2FL

Base

Sum

95.3

91.2

89.0

88.5

238.4

RC

Non-Structural
Steel

MachineWhole Building [t]

602.4

16

Table 2.2

Design material properties.

(a) Concrete(b) Steel Bar

σB

2(N/mm )σB

Grade

Anormal(mm2)(N/mm )

σB
(N/mm2)2Foundation

33

D22

SD345

387

345

490

Upper Part

27

D19
D13

SD345
SD295

287
127

345
295

490
440

D10

SD295

71

295

440

D10

KSS785

71

785

930

4@3 m
z

Shaking Tablezx

Shaking Table

Conventional

Figure 2.4

Figure 2.5

Post‐tensioned

Elevation view of specimens.

Reinforcement stress-strain relations.

17yFigure 2.6

Concrete stress-strain relations.

2.2.1 Japanese Standard Law Provisions
The RC buildings were designed to conform to the Japanese Building Standard Law. The
Japanese seismic design procedure consists of two stages design; allowable stress design formoderate earthquake level to guarantee the damage control performance, and lateral load
capacity design for major to rare earthquake to guarantee the collapse prevention performance.
The base shear coefficient Cb for the allowable stress design is 0.20. The lateral force
distribution shape is an Ai distribution, which is similar to inverted triangular where thelateral load at the`top-most stories is slightly larger. For the structural analysis, the building
was modeled as linearly elastic. All member response was designed to not exceed the yielding
level for reinforcing bars, and the concrete stress response was designed to not exceed theallowable compressive stress of concrete—two third of concrete design strength.
The design base shear coefficients Cb for the lateral load capacity at collapse
mechanism of the conventional RC Building were 0.30 in the frame direction and 0.35 in thewall-frame direction, respectively, as all structural members were designed to perform at the18highest possible ductility. The lateral capacity of the building was confirmed by pushover
analysis that considered nonlinear material characteristics; the lateral force distribution shape
Ai was used. Capacity design checks were carried out for shear failure of beams, columns, andshear walls, as well as shear failure of beam-column joints; note that there was no requirement
regarding the column-to-beam strength ratio at the beam-column joints. Shear reinforcement
provided in columns and beams (in the moment frame or x-direction) and walls (in the ydirection) had shear reinforcement in excess of that required by the Japanese BuildingStandard Law. Minimum requirements such as the spacing of the steel, anchorage detail,dimension of concrete section as well as concrete cover thickness were designed inaccordance with the AIJ Standard for reinforced concrete structures. Thus the RC Building
accurately represented a building that followed typical construction practices common in
Japan.
2.2.2 Assessment of RC Building using ASCE 7-05 and ACI 318-08
A detailed assessment of the RC Building was conducted to assess whether the final design
satisfied U.S. code provisions. This assessment is covered in two subsections—one for the
shear wall direction and one for the moment frame direction—to provide the reader withinformation to help understand the measured responses and observed behavior once thisinformation becomes available.
2.2.2.1 Shear Wall Direction

For the shear wall (y-) direction, the structural system was assumed to be a Building Frame
System Special RC Shear Wall ( R = 6, Cd = 5 ) as the framing provided by the shallow beam
and column at the building edge was insufficient for a Dual System designation. Based on thisdesignation, all lateral forces are resisted by the shear wall. Given that the building system isrelatively simple, the ASCE 7-05 S12.8 Equivalent (Static) Lateral Force Procedure was used,assuming that the building was located in a region where the mapped short period and 1-secperiod accelerations were 1.5 and 0.9, respectively; for Site Class B, design spectral
acceleration parameters were 1.0 and 0.6 with T0 = 12 and TS = 0.6 .
The seismic weight (ASCE 7-05, 12.7.2) of the building was taken as the combined
dead and live loads as 3630 kN (see Table 2.1), i.e., the live load value includes permanent
19

live load attached to the building. The fundamental period of the building was computed using
a two-dimensional model of a single wall, i.e., a cantilever assuming an effective moment ofinertia I eff = 0.51g over the full wall height and one-half the seismic weight at the floor levels.

A fundamental period of T = 0.58 sec was computed from an eigenvalue analysis. According

to ASCE 7-05 12.8.2, Ta = 0.488 ( hn = 12 m )0.75= 0.315 sec Ta and Tu = CuTa = 1.4Ta = 0.0440 ;therefore, T = 0.44 = Tu was used to determine a base shear of V = Cs = 0.167W = 302.5 kN .
Because only two shear walls were used—one at each end of the building—the redundancy
factor

(ASCE

7-05

12.3.4)

was

taken

as

1.3.

Therefore,

Eh = ρ QE = 1.3 ( 302.5 kN ) = 393.3 kN (ASCE 7-05 Equation 12.4-3). Vertical earthquake

loading ( EV ) was included in the load combinations (ASCE 7-05 12.4.2 and 12.4.2.3).
Strength Requirements for Walls: Dead and live loads for the wall were calculated

by assuming the dead and live loads (see Table 2.1) were uniformly distributed based on atributary area equal to the wall length (2.5 m) plus the beam clear length (2.5 m + 2.1 m) times
one-half the joist spacing and the slab overhand (0.9 m + 0.8 m), or 7.82 m2 (84.2 ft2). Shownin Figure 2.7, the resulting story forces produce wall base moment M u = 3569 kN-m andaxialload Pu = 285 kN .

Note

that

the

axial

load

ratio

is

low

⎡⎣ Pu Ag f c′ = 285 kN ( 0.25 m × 2.5 m )( 27 MPa ) = 0.017 ⎤⎦ . Demands were compared with awall P-M interaction diagram (see Figure 2.8), demonstrating that the wall P-M strength does
not satisfy ASCE 7-05 12.8 requirements.
Capacity(

DesignChecks:

Wall

shear

strengthwas

computedas)

φVn = 0.75 Acv α c f c' + ρt f y = 912 kN , using the minimum horizontal web reinforcing ratio(2D10 @ 200 mm spacing for the wall at axis C, tw = 250 mm ; ρt = 0.0031 ; α c = 0.167 ;f c' = 27 MPa ; f y = 345 MPa ). Calculated shear strength φVn = 912 kN is much greater thanshear demand Vu = 393 kN , as would be expected given the relatively high wall aspect ratio(12 m 2.5 m = 4.8 ) . The wall shear strength at axis A is much larger as a result of the 125 mm
spacing of the horizontal web reinforcement.

20161.6
117.7
76.2
37.8

z

y

Vbase = 0.22W = 393.3 kN
Figure 2.7

Equivalent lateral loads on the shear wall system.

Figure 2.8

P-M interaction diagram for the wall.

21

Lateral displacements and story drifts were

Drift Requirements in the Wall:computed according to ASCE 7-05 12.8.6 and compared to allowable story drift per Table
12.12-1 where 0.02hsx ρ = 1.3 = 0.0154 hsx . Story drift ratios of 0.0045, 0.0113, 0.0151, and
0.0167 were computed (Figure 2.9). The drift ratio for the fourth level exceeded the ASCE 705 limit by 8% (0.0167/0.0154 = 1.08).
Detailing Requirements in the Wall: Detailing requirements at wall boundaries werechecked using the displacement-based approach of ACI 318-08 21.9.6 (21.9.6.2); the roof drift
ratio (δ u hw = 0.142 /12 m = 0.012 ) exceeded the minimum value of 0.007. Based on thisvalue, the critical neutral axis depth using ACI 318-08 equation (21-8) is 352 mm. The neutral
axis depth computed for the given wall cross section for an extreme fiber compression strainof 0.003 with Pu = 285 kN is 244 mm; therefore, special boundary elements are not requiredper21.9.6.2.

Thevertical[ ρ = 6 Ab h(2 x + a ) = 0.017 ,reinforcing

with

ratio

of

Ab = 284 mm 2 ,theboundary

reinforcement( 2 x + a ) = 400 mm ],h = 250 mm ,exceeded ρ = 2.3 f y = 0.0067 , where f y = 345 MPa ; therefore, ACI 318-08 21.9.6.5(a) mustbe satisfied as a hoop spacing cannot exceed 203 mm. The configuration and the spacing usedat the wall boundary satisfies the requirements of 21.9.6.5(a), since the spacing of hoops andcrossties is 80 mm (axis A) and 100mm (axis C), and a hoop and a crosstie are provided (all 6bars are supported) over a depth of almost 400 mm, which significantly exceeds the minimum
depth required from 21.9.6.4(a) of one-half the neutral axis depth (244 mm/2).
If the “stress-based” approach of 21.9.6.3 is used, however, the extreme fiber
compression

stressI g S = 0.26 m3 ;

of

and

f c = M u s + Pu A =11.56 MPa
Ag = 0.625 m 2 )significantly( M u = 3569 kN-m ; Pu = 285 kN ;exceeds

thestresslimit

of

0.2 f c' = 5.4 MPa , with 21.9.6.4 left to be satisfied and requiring special boundary elements.

Based on a wall boundary zone with bcx = 160 mm , bcy = 320 mm , Ashx = 2 Ab , Ashy = 3 Ab ,Ab = 78.5 mm 2 , s = 80 mm (axis A) or 100 mm (axis C), f c' = 27 MPa , and f yt = 345 MPa ,the provided Ash values are 1.39 and 2.09 times that required by ACI 318-08 Equation (21-5)
for 100 mm spacing, satisfying 21.9.6.4. Note that the provided Ash values are only 0.45 and0.34 times that required by ACI 318-08 Equation (21-4).

22

In summary, the RC shear wall generally satisfies ASCE 7-05 and ACI 318-08
requirements for the assumed design spectrum, although the wall P-M strength does not meetthe requirement and the interstory drift ratio in the top floor exceeds the limiting value by 8%.
(see Figure 2.9).
2.2.2.2 Frame Direction

For the frame (x-) direction, the structural system was assumed to be a Special Reinforced
Concrete Moment Frame ( R = 8, Cd = 5.5 ) , whereby the lateral forces are resisted by a fourstory, two-bay frame at the perimeter of the building.

Figure 2.9

Interstory drift demands for the wall.

The fundamental period of the building was computed using a two-dimensional modelof a single perimeter moment frame, assuming an effective moment of inertia I eff = 0.3I g forbeams and columns (based on ASCE-41) and one-half the seismic weight at the floor levels. A
fundamental period of T = 0.67 sec was computed from an eigenvalue analysis. According to23

ASCE 7-05 12.8.2, Ta = 00466 ( hn = 12 m )therefore,

T = 0.56 = Tu

was

0.9

used

= 0.44 sec

toand Tu = CuTa = 1.4Ta = 0.610 ;determine

a

base

shear

of

V = CsW = 0.125W = 226.9 kN . The redundancy factor (ASCE 7-05 12.3.4) was taken as 1.3,
since the structure was expected to have an extreme torsional irregularity by loss of moment
resistance at the beam-to-column connections at both ends of a single beam (which is theworst case scenario); therefore, Eh = ρQE = 1.3(226.9 kN) = 294.9 kN (ASCE 7-05 Equation
12.4-3). Vertical earthquake loading ( E y ) was included in the load combinations (ASCE 7-05
12.4.2 and 12.4.2.3).
Strength Requirements for Beams and Columns: Dead and live loads for the beams

and columns—calculated by assuming the dead and live loads (see Table 2.1)—were
uniformly distributed based on a tributary area associated with the member, e.g., for the cornercolumn this is equal to approximately one-eighth the entire floor plan minus one-half the wall
tributary area, or 18.1 m2 (81 ft2) (see Figure 2.10). Using the same spectral acceleration
parameters and seismic weight that were used in the shear wall system calculations, the ASCE
7-05 S12.8 Equivalent (Static) Lateral Force Procedure was used; the resulting story forces areshown in Figure 2.11. These forces were applied to the two-dimensional model to compute themember demands. At the base of the first story, columns values were computed to be Mu =205 kN-m and axial load Pu = 772 kN for the corner columns (C1), and Mu = 200 kN-m and
Pu = 1222 kN for the interior column (C2). Note that the axial load ratio was

Pu Ag f c' = Pu = 772 kN ( 0.5 m × 0.5 m )( 27 MPa ) = 0.11 for the corner columns and 0.18 forthe interior column.

242.7 m
C1

C

G2

G2

C1

W al l

4.4 m

5m
C2

G3

B

Figure 2.10G1G1

7 ,2 0 0

2.7 m
B1

C2

Tributary area for corner column C1.

121.2
88.3
57.1
28.3

z

y

Vbase = 0.16W = 295 kN
Figure 2.11

Equivalent lateral loads on the frame system.

Beam and column nominal moment capacities were computed, and the column, beam,and joint shear demands computed to assess if the system satisfied capacity design concepts
that promote beam yielding. Slab effective widths were based on the provisions of ACI 31808 8.12. Calculation details are provided in Appendix B. The concrete stress-strain relation
was assumed to have a peak of 27 MPa (3.9 ksi) at 0.002 strain, and the steel stress-strain
relation was assumed as an elastic-perfectly plastic behavior with a yield strength of 345 MPa(50 ksi) and an ultimate strength of 490 MPa (71 ksi). Moment and axial load demands of thecolumns were compared with a column P-M interaction diagram (Figure 2.12) and for the25corner column (C1) (Figure 2.13) and the interior column (C2), respectively. The results
demonstrate that the column P-M strengths satisfy ASCE 7-05 12.8 requirements.
In addition, beam moment demands were checked in accordance with the provisions of
ACI 318-08 S21.5 such that M + n > M − n 2 , and neither negative or positive moment strength
at any section along the member length was less than one-fourth the maximum moment
strength at the face of either joint. The amount of reinforcement in the beams
was As , provided = 1140 mm2 ( ρ provided = 0.007 ), which is much greater than the minimum
required reinforcement per ACI 318-08 S21.5.2, As ,min = 654 mm 2 , and less than the maximum
allowed reinforcement ratio ρmax = 0.025 . The reinforcement was continuous along the entire
span, indicating that beam moment strengths satisfy the provisions of ACI 318-08 21.5.

Figure 2.12

P-M interaction diagram for corner column C1.

26

Figure 2.13

P-M interaction diagram for interior column C2.

Capacity Design Checks

Columns Shear Strength (21.6.5): Beam shear demands were determined as whenbeam probable moment strength was reached (calculated using fs = 1.25 fy), column shear
when column probable moments were reached, and beam probable moments reached for theinterior, first-story column [see Figure 2.14(a)] and a typical beam [Figure 2.14(b)]. Nominal
shear strengths also are shown, demonstrating that beam and column shear strengths were
sufficient to develop the beam probable moments, and the column shear strength was
sufficient to resist the column shear developed at column probable moments.
Beam Shear Strength (21.5.4): ACI 318-08 requires that beams of special moment
frames be designed such that flexural yielding occurs prior to shear failure. Therefore, beam
shear strengths were checked to sufficient capacity to resist the shear that develops when the
beam reaches its probable moment of flexural capacity at each end (see Figure 2.15). The
demand calculation was based on the gravity loading on the beams and beam probable
moments. Shear demand and capacity in the beams are also shown in Figure 2.15. Results of27

this assessment are shown in Figure 2.13, demonstrating that beam shear strength satisfied
ACI 318-08 requirements for a special moment frame.
Strong-Column Weak Beam (21.6.2): The strong column–weak beam provision of
ACI 318-08 was checked at all floor levels; this requires that sum of column nominal moment
strength ∑Mnc be at least 1.2 times the sum of the beam nominal moment strengths ∑Mnb.
Column flexural strengths were calculated with the factored axial force, resulting in the lowest
strength [where (0.9-0.2SDS) D + ρE)]. Beam nominal strengths were calculated including aneffective slab width per ACI 318-08 8.12. Results presented in Figure 2.16 demonstrate that
corner columns satisfy these requirements, whereas interior columns have the column-to-beamstrength ratios about 1.0 (< 1.2). Note that the ratio at the roof level connections is smaller
than 1.0, indicating that column yielding might occur at the roof level.
The design of beam-column joints was calculated according to ACI 318-08, Section
21.7, defined as: (1) joint shear demand Vu; (2) joint nominal shear strength φVn; (3) required
transverse reinforcement; and (4) required anchorage. Next, each of these parameters areassessed to determine whether or not the given requirements are satisfied for an interior
connection (case 1: G1-C2-G1), and for an exterior connection (case 2: G1-C2). Additional
details and information for other connections are provided in Appendix B.

M pr ,col = 1.25( M n ,col = 486) = 607 kNm

M + pr ,b = 1.25( M + n,b = 386) = 483kNm

M − pr ,b = 715kNm

M pr ,col = 1.25( M n ,col = 486) = 607 kNm

VE
VE

(2)

=

(1)

=

2M pr ,colh=2(607kNm)= 506kNh = 2.4m

M + pr ,b + M − pr ,b 1198kNm== 499kNhh = 2.4m

φVN = (0.75)(VN = 675) = 506kN

Figure 2.14

Column shear strength demands.

28

w = 10.4kN / m

M + pr ,b = 482.1kNm

M − pr ,b = 715.3kNm⎡⎣Vu , pr ⎤⎦ = ⎡⎣Vu , pr ⎤⎦ =Rmax

M + pr ,b + M − pr ,b wg l+
= 214kN2
l

φVN = (0.75)(VN = 289) = 217kN

Figure 2.15

Beam shear strength demands.

0.70

0.50

0.90

1.44

1.04

1.83

1.25

1.00

1.74

1.21

1.01

1.79

Figure 2.16

Column-to-beam strength ratios.

.

29

Given the weak-beam requirements and capacity design requirements for beam andcolumn shear, beams that frame into beam-column joints are typically assumed to yield prior
to the columns. Therefore, the demands on the joint are controlled by the quantity of
longitudinal reinforcement used in the beams, as well as the stress developed in these bars. In
ACI 318-08 S21.5.4, the probable moment is calculated for a minimum longitudinal
reinforcement stress of 1.25fy. Joint shear demand for both cases was calculated using
horizontal joint equilibrium (Figure 2.17) resulting in: Vu,joint,1 = 1.25As,b1fy + 1.25As,b2fy-Vc1 foran interior connection (case 1), and Vu,joint,2 = 1.25As,b2fy-Vc1 for an exterior connection (case2). Here, Vc1 represents the column shear, which can be estimated as Vc1 = Mc1/(hclear/2) whereMc1 = Mc2 ≈ (Mpr,b1 + Mpr,b2)/2 for case 1, and Mc1 = Mpr,b1/2 for case 2. According to Section
21.7.4, joint shear demands for case 1 and case 2 are Vc1,1 = 936 kN and Vc1,2 = 538 kN,
respectively. Using values of φv = 0.85, and γv = 12 (for both cases), the joint shear capacities
calculated according to Section 21.7.4 are: φVu,1 = φVu,2 = 1097 kN. Note that the nominal
shear capacities are greater than shear demands.

(a)

(b)

Figure 2.17 Free body diagrams for (a) interior and (b) exterior beamcolumn connection.

30

Figure 2.18 Interstory drift demands for the frame system.

Drift Requirements in the Frame: Lateral displacements and story drifts werecomputed according to ASCE 7-05 12.8.6 and compared to allowable story drift per ASCE 705 Table 12.12-1 of 0.02hsx / ρ = 1.3 = 0.0154hsx . As was done to determine the fundamental
period, effective moment of inertia values of 0.3Ig were used for the beams and columns basedon ASCE 41-06 recommendations. Story drift ratios of 0.0099, 0.0134, 0.0108, and 0.0068
were computed, and, the drift ratios did not exceed the ASCE 7-05 limit (Figure 2.18).
Detailing Requirements: Detailing requirements for columns were compared with

ACI 318-08 S21.6.4 provisions. Spacing of the transverse reinforcement in the columns wascompared with the ACI 318-08 S21.6.4.3 provisions where the minimum required transverse
reinforcement spacing is:smin = min(h / 4 = 125 mm; 6dlb = 132 mm; so = 140 mm; 6 in. = 152.4 mm) = 125 mm
where so = 4 + (14 − hx 3 ) and hx = 240 mm Using ACI 318-08 S21.6.4.4, the minimum
required spacing was also calculated to provide the transverse reinforcement. For example, for31the interior column at the base, transverse reinforcement quantity was obtained as
Ash = 4 Ab = 314 mm 2 ,where

Ab = 78.5 mm 2 ,smin = 73 mm

(ACI

318

21-4)

and

smin =107 mm (ACI 318 21-5), where fc = 27 MPa, fy = 345 MPa, bc = 417 mm ,Ag = 250, 000 mm 2 , and Ach = 417 2 mm2 .
smin =

smin =

Ash= 73 mm⎤f c′ ⎡ Ag
0.3bc ⎢(
− 1) ⎥
f y ⎣ Ach
⎦

Ash

f′
0.09bcfy=107 mmEq. (1) (ACI 318 21-4)

Eq. (2) (ACI 318 21-5)

Therefore, the spacing provided in the column (s = 100 mm) satisfies all spacing
requirements except smin = 73 mm determined from (Eq .21-4). This spacing requirement is
not satisfied either at the other floors or in the corner columns. Note that the required
transverse reinforcement should be based on these limits within a height of lo, which islo = min (h = 500 mm; 1/6hclear = 400 mm; 18 in. = 152.4 mm) = 400 mm (see Figure 2.19).
Beyond lo, ACI 318 limits the spacing tosmin = min (6dlb = 132 mm; 6 in. = 152.4 mm) = 132 mm
therefore, beyond lo (i.e., within the middle portion of the column height), ACI 318
requirements are satisfied because s = 100 mm is used.
Detailing requirements at the beams also were checked using ACI 318-08 S21.5.3.
Hoops are required over a length equal to twice member depth (2h region = 1200 mm) (seeFigure 2.19). Minimum required spacing in this region was calculated assmin = min (d / 4 = 150 mm; 8dbl =176; 24dbt =240;12 in. = 304.8 mm) = 150 mm
which does not satisfy the provision, since the provided spacing is s = 200 mm. Beyond the 2hregion, where hoops are not required by ACI 318, minimum spacing is defined assmin = d / 2 = 273 mm and is satisfied.

32

Required transverse reinforcement in the beam-column joints is calculated according
to Section 21.7.3.1. Since bw < ¾ bcol, the required transverse reinforcement is 100% of Ashcomputed for columns. This provision is not satisfied for the same reason as found in the caseof columns (see detailed discussion in the previous section regarding this issue). Development
length of bars in tension was calculated according to Section 21.7.5 [ldh = fy db / (65 (f'c)0.5 ].
For both cases of joints this provision is satisfied since the actual development length isgreater than the required value.

RG1

RFl.

3,000

4C1

h

2h region

3,000

4Fl.

Lo = 40 cm 4G1
3C1

Figure 2.19 Locations where special hoop requirements are needed.

2.2.2.3 Collapse Mechanism

A collapse mechanism analysis was conducted for both the shear wall and moment frame
directions using the code prescribed distribution of lateral forces over the building height.
Four different collapse mechanisms were assumed for each direction: column yielding at thefirst, the second, the third, and the fourth floors. Figure 2.20 shows base shear calculated foreach collapse mechanism assumption. For the moment frame, the expected collapse
mechanism is beam hinging accompanied by hinging at the base of first floor columns and at
the top of the second floor columns (Figure 2.21). For the shear wall direction, the mechanism
involves beam hinging accompanied by yielding at the base of first floor walls (Figure 2.22).
The actual strength coefficients are approximately 0.45 and 0.50 for the moment frame and33

wall-frame directions, respectively, or 3.6 and 3.0 times the values given in ASCE 7-05. Notethat the overstrength factors given in ASCE 7-05 Table 12.2-2 are 3.0 and 2.5 for the moment
frame and shear wall, respectively. Therefore, the computed overstrengths for the wall andmoment frame are higher than expected (3.6 versus 3.0 for frame and 3.0 versus 2.5 for shear
wall direction).

Figure 2.20 Collapse mechanism assessment-influence of column yielding level.

Vbase = 819 kN
Figure 2.21

Controlling collapse mechanism in the frame direction.

34

Vbase = 905 kN
Figure 2.22 Controlling collapse mechanism in the wall direction.

2.3

POST-TENSIONED BUILDINGS

Table 2.3 details the weight and material properties of the specimen. The weight of each floor
from the second to the fourth floor was about 900 kN and the weight of roof floor was 1000
kN. The weight above the foundation was about 3700 kN. The design strength of the precast
concrete was 60 N/mm2. The plan is shown in Figure 2.3 and the elevation in Figure 2.4. The
columns were 450 mm x 450 mm square, the walls 250 mm x 2500 mm thick, and the beams
300 mm x 500 mm in the longitudinal direction. The beam of interior frame was 300 mm x
300 mm in the transverse direction, and the beam of exterior frame was 300 mm x 300 mm.
The floor slab was 130 mm thick. Beams 300 x 300 mm square supported the floor slab atintervals of 1.0 m in the transverse direction.

35

Table 2. Design material properties of post-tensioned specimen.

σy

σt

Grade

Anormal(mm2)(N/mm2)(N/mm2)

SD345

387

385

563

PT bar φ21 (1-3Fl column)*

C

346.4

1198

1281

PT bar φ21 (3-RFl column)*

C

346.4

1189

1273

Grade

Anormal
(mm2)

Fy
(kN)

Ft
(kN)

PT wire φ15.2 (ED of wall base)*

140.7

250

277

PT wire φ15.2 (beam)*

140.7

255

279

PT wire φ17.8 (beam)*

208.4

356

404

PT wire φ19.3 (beam)*

243.7

429

481

D22 (ED for wall base)

STEEL*σy of 0.2% offset

CONCRETE* Fy of 0.2% offsetσΒ

Fc(N/mm2)(N/mm2)Precast concrete (normal)

60

83.2

Precast concrete (fiber)

60

85.5

Top concrete

30

40.9

GROUT

Grout
σΒ

Fc
(N/mm2)(N/mm2)

Column base, wall base and beam end

60

135.6

Wall base (fiber)

60

120.3

PT duct of PT bar and PT wire

30

63.4

The specimen was designed with a typical Japanese PT frame structure in thelongitudinal direction, but with a new type of unbonded PT wall-frame structure in the36transverse direction. Table 2.4 lists the reinforcing details. Figure 2.23 shows details of thewhole steel arrangement. Beam to column connection detail, details of wall, and theconstruction procedure are provided in Appendix A.3. The precast concrete members wereassembled at the construction site, and then half-precast beams and half-precast slabs were
fixed using topping concrete. The half-precast slabs were supported by pretensioned,prestressed beams at 1-m intervals. The design strength of the topping concrete was 30
N/mm2. The design strength of the grout mortar was 60 N/mm2. The PT reinforcement of thecolumns was a high-strength steel bar whose nominal strength was 1080 N/mm2. The PT
reinforcement of beams and walls was high-strength steel strands whose nominal strength was
about 1600 N/mm2. The PT tendons located in sheaths of columns and beams of thelongitudinal direction were grouted. The PT tendons located in sheaths of walls and beams in
the transverse direction were not grouted and remained unbonded from anchor to anchor. The
normal steel bars cross the wall and foundation interface remained unbonded in half of thefirst story wall length. The nominal strength of the normal steel bar was 345 N/mm2. The
column, wall, and beam of the longitudinal direction contained the amount of shear
reinforcement required by the Japanese Building Standard Law. In the transverse direction,
the walls and beams were confined by high-strength steel bars. The nominal strength of thesteel bar was 785 N/mm2. In the first and second stories, one of two walls was additionallyreinforced by steel fibers.
The corresponding grout beds were reinforced by steel fibers as well. The steel fiber
for the wall concrete was 30 mm long with a nominal strength of 1000 N/mm2. The steel fiberfor grout bed was 10 mm long with a nominal strength of 1500 N//mm2. The effective stressof the PT tendon was designed to be 0.6 times of the yield strength for the walls and beams in
the exterior frame of the transverse direction. The effective stress of the PT tendon was
designed to be 0.8 times of the yield strength for the others.

37

Table 2.4 Reinforcement details for PT building.
List of Wall

List of Column

P WallPC1
D13@100
125
4Fl.
3Fl.
2Fl.

450
TendonRebar
Hoop

8-21mm(SBPR1080/1230)
4-D19
D10@100

250

SectionSection

125

450

4Fl.
3Fl.
2Fl.
1Fl.

869

250

480

4C-1-19.3mm(SWPR7BL)

Top
BottomStirrupWeb

2 - D19
3 - D19
2-D10@100
2-D10@200

125
125

250

BottomSection

1,200
381 3812,500
D13(KSS785)@75
D13(KSS785)@75
4-D13
D13@754-D22

1Fl.

400 100

120100 280

125

400 100

120 380

400
100

Tendon480

5-D13@120
Tendon
V bar
H bar

1,200
381 3812,500

3-10-15.2mm(SWPR7B)

List of Girder
PG3

PG2

300

Tendon4C-3-15.2mm(SWPR7BL)

Top
BottomStirrupWeb

2 - D19
3 - D19
2-D10@90
2-D10@200

List of SlabTop Concrete

Section
All

300

300

Top Concrete

400 100

120100 280

Location
400 100

Section3Fl.
2Fl.

400 100

2 - D19
3 - D19
2-D10@200
2-D10@150
500
120 260 120

4C-1-15.2mm(SWPR7BL)

300

Top
Section

5-D13@120

Top
BottomStirrupWeb
Section
4Fl.

300

Tendon500
120 260 120

RFl.

500
120 260 120

SectionCenterTop Concrete869

D13(KSS785)@75
D13(KSS785)@75
4-D13
D13@75

125

PG1
End

2,250
381 3812,500

3-10-15.2mm(SWPR7B)
D13@150(double)
D13@100(double)

TendonV bar
H bar

List of Girder
Location376

D13@100

300

300

Tendon 2C-1-17.8mm(SWPR19L)
Top
BottomStirrupWeb

2 - D192 - D19
2-D10@100(KSS785)

2C-1-17.8mm(SWPR19L)
1C-17mm(SBPR930/1080)
2 - D192 - D19
2-D10@150

Depth: 130mm

Shorter direction Longer direction

CS2
CS3

D10@200

D13@200
D10@200
D10@200
D10@200
D13@200
D10@200

D10@250
D10@250
D10@250
D10@250
D13@200
D10@200

CenterTendon2c-15.2mm

38

End

2,000

Top Concrete

100
30

CS1

Top
BottomTop
BottomTop
Bottom

D10@200

300

PS1

139.1
Half Precast Panel

450 25 2,075
6,700

RPG1

3,000
3,000

2PC1
Tendon2C-3-15.2mm
2PG1

3PC1
3PG3

Tendon1-10-15.2mm
2PC1
PCBar2-21mm
PCBar3C-21mm

200 100
300

170
130

PW25
PCBar2-21mm
PCBar3C-21mm

Tendon2C-1-17.8mm

2PG2

2Fl.

4PG3

Tendon2C-1-17.8mm

3PG2

2PG3

FG2

4-D25
FG3

2-D13
2-D13
4-D25

7,200

7,200
14,400

A

3,600

3,600
7,200

C

B

4-D25
3-D25

850

Mortar1Fl.

1,500
Bondless

FG1

1PC1
2-D22

850

Tendon2C-9-12.7mm
FG1
6-D25

1PC1

1,200

1,020 180

1PC1

1Fl.

Tendon2C-1-17.8mm

PW25
Coupler

3Fl.

250

4PC1

PW25

Mortar
1PC1
CouplerMortar

30

2PG1

4PG2

4Fl.

2PC1
6,370

2Fl.

850

6,370

2PC1

3,000

3PG1

Tendon2C-1-17.8mm

PW25

3PC1

30

3,000

400

3Fl.

3PC1
Tendon2C-3-15.2mm
3PG1

12,000100

PCBar3C-21mm30

12,000PCBar2-21mm

4PG1

3,000

3,000

4Fl.

400

RPG3

Mortar

4PC1

3,000

5,970
500
260
120 120

Top Concrete
Tendon2C-1-19.3mm
4PG1

5,970

4PC1

280
120
100

3,000

150

1,200
100 1,100
500600 30

RPG2

RFl.
150

4PC1

45025

450

RFl.

25

1,000

225 225
Tendon2C-1-15.2mm
RPG1

2,500

870 760 870

400

400

225 2253PC1

25

25 450300 450 1,000

450 25

3

3a

7,200

4

3-D25
4-D25
4

3

Figure 2.23 Configuration of the steel.

In designing the columns and beams in the longitudinal direction, more than 1.5 of thecolumn-to-beam strength ratios was satisfied so that the complete mechanism was based onbeam hinges. The strength capacity in the longitudinal direction was set to have the samevalue as defined in the Japanese Building Standard Law. The PT wall was designed referringto static parametric studies using a fiber model. The study focused primarily on the balance
between the amounts of vertical PT tendons and the confinement reinforcements, as well as on
the influence to capacity of the normal unbonded steel bars of the base. Basically, the walls
satisfied the provisions of ACI ITG-5.2-09. Detailed information of unbounded post-tensioned
concrete walls was as follows:
Unbonded Post-Tensioned Concrete Walls:

The four-story unbounded post-

tensioned (UPT) concrete walls were constructed using four precast concrete panels that were39

post-tensioned together along horizontal joints. The typical section for the wall panels was 2.5
m long by 250 mm thick with a cross-sectional aspect ratio ( l w / t w ) of 10. The first, second,
and third story wall panels were 3 m high. The fourth story wall panel was extended 450 mmabove the roof slab. The extended length of the fourth story wall panel was thickened to 400
mm in order to accommodate anchorage for the post-tensioning reinforcing. The assembled
walls had a height-to-length aspect ratio ( H w / l w ) of 5.
The concrete panels for the North wall were fabricated using a high-performance fiber
reinforced cement composite (FRCC). The South wall panels were fabricated using a
conventional Portland cement concrete mix with a minimum specified compressive strength of
60 MPa (8.7 ksi). The vertical faces of the panels were reinforced with a two-way mesh of
D13 SD295 reinforcing bars. Supplemental D13 SD295 transverse ties were added to prevent
separation of the reinforcing mesh from the concrete core, a failure mechanism noted by Perezet al. [2004c]. The mild steel reinforcing was not developed across the panel joints.
The compression zones of the wall panels were reinforced with high-strength S13
KSS785 confinement hoops. In the base wall panel the compression zones were reinforced
with two bundled, overlapping S13 KSS785 hoops at a vertical spacing of 75 mm. The
confinement reinforcing ratios for the base wall panel, equal to the volumetric ratio ofconfinement reinforcing to the confined concrete core, were 1.7% for the length-wise
direction ( ρ x ), and 1.8% for the thickness direction ( ρ y ). The overall confinement reinforcing
ratio ( ρ s ) for the base wall panel was 3.5%. In the upper story panels, the level ofconfinement was reduced to single S13 KSS785 hoop at 100 mm vertical spacing. The ratio of
the total confinement length to the overall length of the wall ( lc / l w ) was 0.4.
Based on preliminary design results presented at planning meetings at PEER, a wallcross section 250 mm thick and 2500 mm long was selected. According to the AIJ Guidelines,
the walls have deformation capacity of more than 2% drift angle for both shear failure andbending compression failure. In the PT Building, the wall was post-tensioned by unbonded
strands extending over the full height of the building to provide a mechanism for energy
dissipation at the interface of the wall and the foundation. Unbonded reinforcement also wasplaced across the interface of the wall base and the foundation to provide a mechanism forenergy dissipation. The arrangement of the unbonded energy-dissipating reinforcement was40

selected based on numerical studies. These studies are briefly described in the following
paragraph.
For both the RC and PT Buildings, preliminary analyses were conducted using fibermodels to assist with design decisions. Two results are presented for the PT Building, onewith two PT strands and no energy-dissipating bars, and the other with two PT strands and 8energy-dissipating bars (Figure 2.24). Figure 2.25 compares relative strength, hystereticenergy dissipation, and concrete compressive strain for RC and PT walls. The energy
dissipation capacity of the PT wall increased four times by providing the unbonded deformed
reinforcement at the wall base (and embedded into the foundation). The concrete compressive
strain was about four times higher in the PT wall compared with the RC wall. In addition toproviding high-strength transverse reinforcement, as was done in the RC wall, steel-fiber
reinforced concrete was used over the first two stories of the PT wall.
In order to enhance energy dissipation during seismic response, eight D22 SD345 mild
steel reinforcing bars (four at each end) were included across the base panel-foundationinterface. The energy-dissipating reinforcing bars were positioned within the central core of
the wall (i.e., outside of the compression regions) and were unbonded over a length of 1.5 m
within the base wall panel. In order to facilitate construction, the energy-dissipating bars werespliced within the foundation using a grouted coupler.
The post-tensioning in the walls consisted of two bundles of 10-D15.2 SWPR7B post
tensioning strands, with a PT steel ratio ( ρ pt ) of 0.44%. The bundled strand groups werepositioned symmetrically on either side of the centroidal axis of the wall with an eccentricityof 380 mm. The initial prestress (after release) in the strand groups was equal to 60% of theyield stress for the strand material ( f py ) . The corresponding initial compressive stress in thewall due to post-tensioning ( f ci , pt ) was 4.3 MPa (0.62 ksi). Because the bundled strands were
contained within ungrouted polyethylene ducts, they were unbonded from the concrete wall
panels over the full wall height between mechanical anchorages at the top and bottom of the
wall

41

Shear Force (kN)
800
600
400
200
0
‐0.04

‐0.03

‐0.02

‐0.01

‐200

0

0.01

0.02

0.03

0.04

‐400
‐600
‐800

R (rad)(1) RC WallShear Force (kN)Shear Force (kN)
800

800
80

600

600
60

400

400
40

200

200
20
0

0
‐0.04

‐0.03

‐0.02

‐0.01

‐200

0

0.01

0.02

0.03

‐0.04

0.04

‐0.03

‐0.02

‐0.01

0

0.01

0.02

0.03

0.04

‐200
‐20
‐400
‐40

‐400

‐60
‐600

‐600

‐80
‐800

‐800

R (rad)R (rad)(2) PT Wall: 2 PT ducts- 8 energy dissipating bars

(3) PT Wall: 2 PT ducts- no energy dissipating bars

Figure 2.24 Hysteretic behavior of cantilever analyses.
Hysteretic area‐Energy consumption (kN*m)

Lateral force capacity(kN)
500

250

400

200

300

150

200

100

100

50

0

0

PT 2‐8PT 2‐0RC

PT 2‐8PT 2‐0Compressive strain of the extreme fiber (%)
1
0.8
0.6
0.4
0.2
0
PT 2‐8PT 2‐0RC

Figure 2.25 Strength, hysteresis, energy dissipation, and concrete
compressive strain at 2% drift angle.

42RC

2.3.1 Design of Unbonded Post-tensioned Concrete Walls
2.3.1.1

Performance-Based Design

Details for the UPT concrete walls were developed using a performance-based design
approach. For design purposes, the UPT concrete walls were conservatively analyzed as
isolated lateral force resisting components, i.e., the contribution of the light PT frames and theinteraction of the walls with the connecting UPT beams and composite floor system were
neglected. Two analytical models were developed to characterize the lateral load response of
the walls and to estimate design capacities and design demands: (1) an idealized tri-linear
lateral load response model; and (2) a rigorous nonlinear finite element model (presented in
Section 2.3).
Idealized Tri-Linear Lateral Load Response Model: Previous analytical andexperimental studies [Kurama et al. 1996; 1997; 1999a; 1999b; Perez et al. 1998; 2004a;
2004b; 2004c; 2007; Keller and Sause 2010] have demonstrated that the lateral load responseof UPT concrete walls can be characterized by the following limit states: (1) decompression(DEC), (2) effective linear limit (ELL), (3) yielding of the post-tensioning steel (LLP), (4)crushing of the confined concrete (CCC), and (5) fracture of the post-tensioning steel (FP).
For well-designed and detailed UPT concrete walls, an idealized tri-linear pushover curve(Figure 2.26) can be developed using simplified predictions of response parameters for limit
states 2 (ELL), 3 (LLP), and 4 (CCC). Comparisons of response predictions from the idealized
tri-linear pushover model with results from previous large-scale experimental tests anddetailed nonlinear finite element analyses are presented in Figure 2.27.

43

Figure 2.26 Idealized tri-linear lateral load response curve for UPT
concrete walls [Perez et al. 2004a].

Figure 2.27 Comparison of experimental and analytical results for test
wall TW5 [Perez et al. 2004a].

44

LIMIT STATES FOR UPT CONCRETE WALLS:
Decompression (DEC)—Decompression (DEC) occurs when tensile strain demand at thebase of the wall, due to overturning moment from lateral loading, equals the pre-compression
strain due to post-tensioning and gravity loads. If reinforcing steel is not developed across thehorizontal joint at the base of the wall, decompression is accompanied by the initiation of gap
opening along the wall base-foundation interface. Under a specified lateral load distribution,
decompression of the wall can be related to a specific level of base shear, Vdec , and roof drift,Θ dec .
Effective Linear Limit (ELL)—The lateral load response of a UPT concrete wall is nearlylinear elastic immediately after decompression. As drift levels increase, however, a substantial
reduction in lateral stiffness occurs due to nonlinear softening of the concrete in compression
and the progression of the gap opening along the horizontal joint at the base of the wall
(geometric softening). The lateral stiffness decreases in a smooth and continuous manner, so
the term effective linear limit is generally used to describe the point at which softening isapparent.The base shear and roof drift corresponding to the effective linear limit are Velland Θ ell , respectively.
Yielding of the Post-Tensioning Steel (LLP)— The linear limit for the post-tensioning steel

is calculated at the onset of yielding. For simplicity, the axial strain demand is calculated at
the centroidal axis of a strand group, i.e., small discrepancies in strain within a group due tothe relative eccentricity of the individual strands are neglected. The LLP limit state for thewall is reached when tensile strain demand in the critically stressed group reaches the linear
limit for the strand material. The base shear and roof drift corresponding to yielding of the
post-tensioning steel are denoted as Vllp and Θ llp , respectively.

Crushing of the Confined Concrete (CCC) —Failure of the wall occurs when the confinedconcrete at the base fails in compression. Based on the confined concrete constitutive model
developed by Mander et al. [1988a; 1988b], crushing of the confined concrete occurs at an
ultimate concrete compressive strain, ε cu , which is reached when the confinement
45

reinforcement fractures. Significant loss of lateral load and gravity load resistance are
expected to occur when the crushing limit state is reached. The base shear and roof drift
corresponding to crushing of the confined concrete are denoted as Vccc and Θ ccc , respectively.
Fracture of the PT Steel (FP)—Fracture of the PT steel occurs when the tensile strain

demand reaches the capacity of the strand material. The limit state is accompanied by asudden and significant loss of lateral load resistance and self-centering capability. The baseshear and roof drift corresponding to fracture of the post-tensioning steel are denoted as V fp
and Θ fp , respectively.

DESIGN CRITERIA FOR UPT CONCRETE WALLS

The following design criteria were developed by Perez et al. [2004c] for UPT concrete walls:
Criterion 1: Softening—This design criterion controls softening of the lateral stiffness of the

UPT concrete wall for the design level ground motion.
Vell ≥ α d ⋅ Vd
where Vell is the base shear at the effective linear limit, α d is a factor applied to the designbase shear demand to define the base shear at which softening is allowed to occur(recommended range: 0.65-1.0), and Vd is the design base shear demand.
Criterion 2: Base Moment Capacity—This design criterion controls the base moment

capacity of the wall as governed by axial-flexural behavior.
Φ f Vllp ≥ Vd

where Φ f is a capacity reduction factor for flexural strength, and Vllp is the base shear
corresponding to the initiation of yielding in the PT steel.
Criterion 3: Yielding of the Post-Tensioning Steel—This design criterion controls yieldingof the PT steel, which has an adverse effect on drift control and self-centering capability.
Θ llp ≥ Θ d46

where Θ llp is the roof drift corresponding to the initiation of yielding in the PT steel and Θ dis the roof drift demand for the design level ground motion.
Criterion 4: Story Drift—This design criterion controls the maximum story drift for thedesign level ground motion.

δ all ≥ δ dwhere δ all is the allowable story drift for the design level ground motion, and δ d is the story
drift demand for the design level ground motion.
Criterion 5: Crushing of the Confined Concrete—This design criterion controls the axial-

flexural compression failure of the walls.
Θ ccc ≥ Θ mwhere Θ ccc is the roof drift corresponding to crushing of the confined concrete, and Θ m is
the roof drift demand for the maximum considered ground motion.
Criterion 6: Fracture of the Post-Tensioning Steel—This design criterion ensures thatfracture of the PT steel does not occur.
Θ fp ≥ Θ ccc

where Θ fp is the roof drift corresponding to fracture of the PT steel.

Criterion 7: Roof Drift Limit under the Maximum Considered Ground Motion—This

design criterion limits the drift demand under the maximum considered ground motion toensure stability of the gravity load system.
Θ g ≥ Θmwhere Θ g is the roof drift corresponding to failure of the gravity load resisting system.

47

ESTIMATION OF DESIGN CAPACITIESPreliminary estimates of design capacities for the walls were based on the simplified tri-linear
lateral load response model.

Perez et al. [2004c] presents simplified expressions for

estimating design capacities of UPT concrete walls. Final estimates of design capacities forthe walls were based on nonlinear finite element pushover analyses (see Section 2.3).
ESTIMATION OF DESIGN DEMANDS

Design demands for the UPT concrete walls were based on three levels of seismic intensity.
Seismic response coefficients ( C s ) of 0.20 and 0.30 were used to represent the design-basis
earthquake (DBE) and the MCE, respectively. In addition, the UPT concrete walls were
designed to remain linear elastic up to a seismic response coefficient of 0.15. Preliminary
estimates of deformation demands for the UPT concrete walls were estimated using theprocedure outlined in Seo and Sause [2005], which accounts for the tangent stiffness of thewall after the effective linear limit (ELL) and hysteretic energy-dissipation characteristics.
Nonlinear response history simulations (see Section 2.3) were used to evaluate deformation
demands for the proposed test plan.
CONFORMANCE WITH CURRENT U.S. DESIGN PROVISIONSThe UPT concrete wall design satisfies the strength and detailing requirements of ACI ITG5.2-09 with one notable exception. The PT reinforcing groups are offset from the centroid of
the wall by 15% of the wall length. The ACI ITG-5.2-09 was developed for UPT concrete
walls with PT reinforcing located within 10% of the wall length from the wall centroid. The
experimental program described herein increased the eccentricity of the PT reinforcing steel to
15% to control drift demands, by way of increasing the post-decompression lateral stiffness.
The two ground acceleration records selected for the experimental program, from the 1995
Great Hanshin Earthquake produce relatively large spectral acceleration demands in theelongated post-ELL period range of the structure, which significantly increases deformation
demands in the structural system.

48

2.4

CONSTRUCTION

The buildings were constructed between July and October 2010 and moved onto the EDefense shake table in November 2010. Instrumentation of building was primarily completed
in November 2010. The construction process is depicted in Appendix C.
The specimen was constructed outside and then transferred onto the shake table, as
shown in Appendix A. The specimen was suspended by two cranes and then set on the
shaking table. The foundation beams were strongly fixed by one hundred and fifty posttensioned PT bars. The foundation beams were constructed on the six concrete stubs, 1.4 m x
3 m x 1.5 m in configuration, to leave enough space for the carrier access beneath thespecimen. The foundation beams were 1200 mm deep and designed for each phase of the test
program, from the construction to set up, by using the supplementary PT tendons to prevent
excessive cracks. The concrete was cast for the columns, walls, upper floor beams, and thefloor slab. The main reinforcement of columns, beams, and the assumed column-zones ofwalls were connected by gas pressure welding. Lap joints were used for reinforcing the walls
and floor slabs.

49

50

3

Test Plan and Instrumentation

The two test buildings were heavily instrumented to assess their performance when subjected
to a range of shaking intensities for a range of post-test analytical studies. The table motions
used for the testing and the instrumentation used for each of the two buildings are briefly
described in the following sections. Additional information is provided in Appendix D.

3.1

TEST PLAN

The 1995 JMA-Kobe and JR-Takatori records were selected for this experimental program.
Testing was conducted on December 13th and December 15th, subjecting the buildings to the
JMA-Kobe record, and a third test was conducted using the JR-Takatori record on December
17th. The NS-direction acceleration, EW-direction acceleration, and vertical-direction
acceleration were aligned with the transverse-direction (y), longitudinal direction (x), andvertical direction of the specimen (Figure 2.3). Natural periods 0.36 and 0.18 were computedfor the models (see Chapter 2) for the shear wall (y) and moment frame (x) directions,
respectively. In the tests the amplitude associated with the JMA-Kobe record was scaled to
produce a range of shaking intensities; scale factors of 25 %, 50 %, and 100 % were used. The
orbit of horizontal acceleration is shown in Figures 3.9-3.10. Based on preliminary analyses,
the stronger NS-direction wave was input into the transverse-direction. The two tests run with
the JR-Takatori record were scaled to 40% and 60%.

51

3.2

INSTRUMENTATION

3.2.1 General

A total of 609 channels of data were collected during the tests for RC and PT specimens,
including 48 accelerometers, 202 displacement transducers, and 235 strain gauges. The
accelerometers were placed on the foundation and on each floor slab to record accelerations inthree directions. Displacement transducers were arranged to measure interstory displacements,
beam end rotations, column end rotations, and base wall rotations. Strain gauges were glued tolongitudinal and transverse reinforcement of beams, columns, and walls. Strain gauges werelargely used for the RC specimen, whereas displacement transducers were used for the PT
specimen (to measure member end rotations). Video cameras were used to record the tests andincluded overall views of the test specimens, as well as close up views of regions where
yielding and damage were anticipated. Data acquisition was accomplished using 24 bit A/Dconverters using a sample rate of 0.001 sec (1000 Hz). Locations of instrumentation are
shown in Appendix D.
3.2.2 Types of instrumentation

Figure 3.1 shows properties of the three different types of instrumentation that were used forthe tests: accelerometers, displacement transducers, and strain gauges.
3.2.2.1 Accelerometers

Accelerometers were used to record accelerations at each floor. Figure 3.2 shows the locationsof accelerometers. Detailed information is provided in Appendix D.

52

Figure 3.1

Properties of the instrumentation used in the specimens.

53

Figure 3.2 Locations of the accelerometers.

3.2.2.2 Displacement Transducers

A total of 202 displacement transducers were used for the tests, including wire potentiometers,
laser-type displacement transducers, and linear variable differential transducers (LVDTs). The
transducers were attached to the test specimens to measure horizontal and verticaldisplacements, lateral story displacements and drifts, average concrete strains over gauge
lengths, pullout/gapping at member ends, and sliding at the base of the shear walls. Locationsof wire and laser transducers are shown in Figures 3.3 and 3.4.
A majority of the LVDTs were provided by NIED; however, some of thedisplacements transducers were provided by NEES@UCLA, IOWA State University, and theEarthquake Research Institute at the University of Tokyo; this enabled more detailed
measurements of wall deformations (Figures 3.5 and 3.6). Four transducers were used over agauge length of 540 mm at the base of the walls to enable the curvature along the wall length(depth) to be determined (Figure 3.5); additional displacement transducers were provided ateach wall boundary over the entire height of the building (Figure 3.5). Two pairs of
diagonally-oriented displacement transducers were used over the first story height to enable
54

the determination of shear deformations. Photographs showing the displacement transducers
over the first story height of the RC building are shown in Figure 3.7. Further information isprovided in Appendix D.
Strain Gauges: Reinforcement strains were measured at 235 locations using strain

gauges. Figure 3.8 shows the locations of the strain gauges in horizontal and verticalreinforcement in RC building at the first and second floor. More detailed information isprovided in Appendix D.

Figure 3.3 Locations of the wire-type displacement transducers.

55

Figure 3.4 Locations of the laser-type displacement transducers.

Figure 3.5

Vertical LVDT configuration (first floor).

56

Figure 3.6 Diagonal LVDT configuration (first floor).

Figure 3.7 Instrumentation on the RC wall.

57

(a)

(b)

Figure 3.8

Strain gauge locations in horizontal and vertical directions
at the first floor (RC).

58

3.3

GROUND MOTIONS

Two different table motions at various intensities were used: JMA-Kobe (25%, 50%, and100%) and Takatori (40% and 60%). The testing was planned over five days: low-to-moderateintensity JMA-Kobe (25% and 50%) on December 13, 2010, 100% JMA-Kobe on December
15, 2010, and Takatori (40% and 60%) on December 17, 2010.
Pseudo acceleration spectra of the JMA-Kobe ground motions are presented in Figures
3.9 and 3.10 for the x- (frame) direction and y- (shear wall) directions, respectively. The
broken lines show the target spectrum, whereas solid lines illustrate the actual spectra
determined form measurements. Peak spectral accelerations observed on the shaking tablewere 0.58g at 25%, 1.18g at 50% and 2.79g at 100% JMA-Kobe in the frame direction; and0.89g at 25%, 1.58g at 50% and 3.42g at 100% JMA-Kobe in the shear wall direction.
Pseudo acceleration spectra of the Takatori ground motions were also plotted (seeFigures 3.11 and 3.12). At 40%, the Takatori record had a peak spectral acceleration of 1.11g
and 0.99g in the frame and shear wall directions, respectively. At 60%, the Takatori record
had a peak spectral acceleration of 1.72g in the frame direction and 1.51g in the shear wall
directions, respectively.
Displacement spectra are shown in Figures 3.13-3.16. Peak spectral displacements
were observed as 10.5 cm at 25%, 20.9 cm at 50%, and 41.8 cm at 100% JMA-Kobe; and 40.3
cm at 40%, and 60.2 cm at the 60% Takatori in the frame direction. In the other direction, thepeak displacements were 11.6 cm at 25%, 23 cm at 50%, and 46 cm at the 100% JMA-Koberecord; and 48.1 cm at the 40%, and 72.3 cm at the 60%Takatori records.

59

Figure 3.9 Acceleration spectra for JMA-Kobe ground motion (x-direction).

Figure 3.10 Acceleration spectra for JMA-Kobe ground motion (y-direction).

60

Figure 3.11 Acceleration spectra for Takatori ground motion (x-direction).

Figure 3.12 Acceleration spectra for Takatori ground motion (y-direction).

61

Figure 3.13 Displacement spectra for the Kobe ground motion (x-direction).

Figure 3.14 Displacement spectra for the Kobe ground motion (y-direction).

62

Figure 3.15 Displacement spectra for the Takatori ground motion (x-direction).

Figure 3.16 Displacement spectra for the Takatori ground motion (y-direction).

63

64

4

Summary, Conclusions, and Future Work

4.1

SUMMARY

Detailed information related to the December 2010 tests of two, full-scale, four-story
buildings that were tested on the NIED E-Defense shake table are presented. Substantial
collaboration between U.S. and Japan researchers over a period of nearly two years preceded
the shake table testing. The goal of the collaboration was to produce test buildings that would
provide vital data on behavior and response over a spectrum on shaking intensities, including
near-collapse, for research efforts in both the U.S. and Japan.
The tests were successfully completed during the week of December 13-17, 2010. The
large number of instruments placed, including video cameras, will provide a wealth of datathat will enable both Japanese and U.S. researchers to improve our understanding of the
behavior of these systems. Papers that summarize the overall results are being prepared forsubmittal to AIJ and a U.S. journal by mid-summer 2011.
Support has been provided by NEEScomm to conduct a blind prediction study
associated with the RC and PT Building tests. The data in this report are intended to provide
background information to support this effort.

4.2

FUTURE STUDIES

A subsequent report will be prepared that provides an overview of the test results and pre-test
analytical studies, as well as post-test studies.

65

66

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69

70

Appendix A

A.1 MATERIAL PROPERTIES
Actual material properties for RC specimen

Steel
σy

σt

Grade

Anormal(mm2)(N/mm2)(N/mm2)

D22

SD345

387

370

555

D19

SD345

287

380

563

D13

SD295

127

372

522

D10

SD295

71

388

513

D10*

SD295

71

448

545

D10*

KSS785

71

952

1055

*σy of 0.2% offset (shear reinforcement)
Concrete
Fc(N/mm2)(N/mm2)σΒAge(Days)

1st - 2nd floor

27

39.6

91

2nd - 3rd floor

27

39.2

79

3rd - 4th floor

27

30.2

65

4th - roof floor

27

41.0

53

71

Actual material properties for PT specimen

Steel
σy

σt

Grade

Anormal(mm2)(N/mm2)(N/mm2)

D22 (ED for wall base)

SD345

387

385

563

PT bar φ21 (1-3Fl column)*

C

346.4

1198

1281

PT bar φ21 (3-RFl column)*

C

346.4

1189

1273

Grade

Anormal
(mm2)

Fy
(kN)

Ft
(kN)

PT wire φ15.2 (ED of wall base)*

140.7

250

277

PT wire φ15.2 (beam)*

140.7

255

279

PT wire φ17.8 (beam)*

208.4

356

404

PT wire φ19.3 (beam)*

243.7

429

481

*σy of 0.2% offset* Fy of 0.2% offset
ConcreteσΒFc(N/mm2)(N/mm2)Precast concrete (normal)

60

83.2

Precast concrete (fiber)

60

85.5

Top concrete

30

40.9

Grout
σΒ

Fc
(N/mm2)(N/mm2)

Column base, wall base and beam end

60

135.6

Wall base (fiber)

60

120.3

PT duct of PT bar and PT wire

30

63.4

72

A.2 MEMBER GEOMETRY AND REINFORCEMENT OF THE RC SPECIMEN

C1

Wall

14,400

G1

B1

Y
1

S1

G2

G2

Wall

C1

C1

2,500
3,600

3,600

1a

3,600
7,200

7,200

X

S1
S1

2,500
3,600

C2

S1

A

Wall

S1

G3 S1

C2

FS1

FS1
A

S1

S1

7,200

7,200

B

B1

G1

7,200

7,200
14,400

B

C1

G2

G2

FS1

G1

C

FS1

G1

Wall

C

1

2

1a

(a) Foundation

2(b) 2-4 stories

Figure A.1 Floor plan of the RC specimen.

3G1

2C1

2C1

2C2
2G1

2Fl.

3C1

3C2
3G1

2G1

1C1

1C1

1C2

1Fl.

7,200

3,000
12,000

4G1

3C1
3Fl.

4C1

4C2
4G1

3,000

4C1
3Fl.

3,000

3,000
3,000

12,0003,000
1,200 3,000

RG1

1,200 3,000

RG1

4Fl.

4Fl.

RG2

3Fl.

4G2

3Fl.

3G2

2Fl.

2G2

Wall

800

A

B

C

(a) 1-frame and 2-frame

Figure A.2

3C1

3C1

Wall

3G3

2C1

Wall

2C1

2G3

2G2

1C1

1C1

1Fl.

3,600

3,600

1,400

7,200

1,400

1

1a

2

(b) A-frame and C-frame

Elevation of the RC specimen.

73

4G3

3G2

3,600

14,400

4C1

4C1
4G2

Wall

7,200

800

RG3

RG2

3,600

500

7,200

500

1

1a

2

(c) B-frameRC specimen

Figure A.3 Overview of the RC specimen.

74

Table A.1 List of steel reinforcement
List of Column

List of Wall
C1

Wall

C2
SectionSection
4Fl.
3Fl.

BxD
Rebar
Hoop
Joint

500 x 500
8-D22
2,2-D10@100
2,2-D10@140

500 x 500
10-D222,2-D10@100
2,2-D10@140

4Fl.
3Fl.
2Fl.

BxD
Rebar

2 x 6-D19

Hoop

2,2-D10@100

Joint

500 x 500
8-D22
2,3-D10@100
2,2-D10@140

500 x 500
10-D22
2,4-D10@100
2,2-D10@140

Section
1Fl.

400
BxD
Rebar

Top
Section

1Fl.

BxD
Rebar
Hoop
Joint

Hoop

CS1
500 x 500
10-D22
3,4-D10@100
2,2-D10@140

500 x 500
10-D22
3,4-D10@100
2,2-D10@140

List of Girder
Location

End

BxD
Top
BottomWeb
Stirrup

BxDTop
BottomWeb
Stirrup

CS3

LocationEnd

D13@300 (W)A D10@125 (W)Horizontal
C D10@200 (W)2,2-D10@150

Depth: 130mm
Shorter direction
D10@200
D10@200
D10,D13@200
D10@200
D10@200
D10@200
D10,D13@200
D10@200

Longer direction
D10@250
D10@250
D10@250
D10@250
D10@250
D10@250
D10,D13@200
D10@200

List of Girder
EndG3

300 x 600
4-D22 3-D22 4-D22
3-D22 3-D22 3-D22
4-D10
2-D10@200

RFl.

BxD
Top
BottomWeb
Stirrup

Location

CenterSection
300 x 300
3-D19
3-D19
2-D19
3-D19
2-D10@100(KSS785)

4Fl.
3Fl.
2Fl.

300 x 600
5-D22 3-D22 5-D22
3-D22 3-D22 3-D22
4-D10
2-D10@200

BxD
Top
BottomWeb
Stirrup

300 x 300
3-D19
4-D19
3-D19
3-D19
2-D10@100(KSS785)

List of beam
B1

BxD
Top

300 x 600
6-D22 3-D22 6-D22

BottomWeb
Stirrup

3-D22 3-D22 3-D22
4-D10
2-D10@200

LocationEnd

CenterSection
All

BxD
Top
BottomWeb
StirrupEnd

CenterSectionBxDTop
BottomWeb
Stirrup

300 x 400
5-D19
3-D19
3-D19
4-D19
2-D10
2-D10@200

SectionSection
4Fl.
3Fl.
2Fl.

Section
2Fl.

CS2

Top
BottomTop
BottomTop
BottomTop
Bottom

2,500 x 250
Vertical

G2

Section3Fl.

2 x 6-D19
2,3-D10@80
2,3-D10@100

List of Girder
G1
CenterSection
RFl.
4Fl.

A
C

List of Slab
S1

Hoop

400

Joint

500 x 500
8-D22
2,3-D10@100
2,2-D10@140

BottomSection
BxD
Rebar
Hoop
Joint

H bar

250

BxD
Rebar
Hoop
Joint

D13@300 (W)A D10@125 (W)Horizontal
C D10@200 (W)2,2-D10@15035d

Section2Fl.

2,500 x 250
Vertical

300 x 400
3-D19
3-D19
4-D19
7-D19
2-D10
2-D10@200

75

1Fl.

BxD
Top
BottomWeb
Stirrup

300 x 400
4-D19
3-D19
3-D19
4-D19
2-D10
2-D10@200

3-D22
4-D22

4-D13@180
8-D25
FG1 8-D25
4-D25

3-D19
1WC25

3-D22
2-D22

3-D22

3-D19

3-D22

3-D22
2-D22

4-D25 4-D25
2-D25 4-D25

4-D19 2G3 3-D19
3-D19

3-D19

1C1

FG2

5-D25
4-D25 FG3 2-D25

1,200

1Fl.

3-D19
2C1

3-D22
2-D22
2G2

3-D19 4-D19

1C1

3-D22

3-D22
2-D22

3-D22

3-D22
2-D22
3-D22
2-D22

12,000FG1

1,200

3,4-D10@100
1Fl.

1C2
4-D22
2-D22

4-D22

4-D22
2-D22

1C1

2Fl.

4-D19 3G3 3-D19

2-D22
3-D22

3-D22

2G1

3-D19
2C1 2WC25

3-D19

2-D22
3-D22

3-D22
3-D22

1,700

3-D22

3,000

3-D22
3-D22

3-D22
2,3-D10@100

3-D22

2G1

3-D22

2C2

3-D19
3C1

3G2

3-D19 4-D19

4-D19

4-D19 4G3 3-D19

D10@150

3-D22

3-D22

3Fl.

2-D22
3-D22

3-D22

3G1

3-D19
3C1 3WC25

3-D22

3-D22
2-D22

4G2

3-D19 4-D19

3-D22
2-D22

3-D22

D10@200
3-D22
3-D22

3,000

3G1

3,000

D10@200

4Fl.

3,000

3-D22

2C1 3-D22

2,3-D10@100
2Fl.

4-D25
8-D25
7,200

2-D25 2-D25
4-D25 4-D25
4-D25
3,600

5,000
14,400

A

B

4-D25
3,600

C

7,200
1

(a) 1-frame and 2-frame

7,200

2

1a

B

1a
7,200

2

C2

G3

C2

cS2

cS2

S1
B1

G1

7,200

S1
B1
S1

cS1

B1
S1
A

C1
cS3

cS1

G2

G2
Wall

cS2

C1
cS3

6,900

(d) Floor slab

Figure A.4 Details of RC specimen.

76

2

1

(c) B-frame(b) A-frame and C-frame
1

2-D25
5-D25

3,600

1,500

4-D25

G1

3,000

3C2
4-D22

3-D22
2-D22

3-D22
2-D22

3,000

12,0002,2-D10@100
3Fl.
2,2-D10@140

3-D22

4-D22

3-D22

3-D22
2-D22

3C1 3-D22

4G1

3-D22

3-D22
1-D22

3-D22

3-D19
4C1

3-D22
2-D22

4G1

2-D19

D10@200

4-D22
2-D22

3-D22
1-D22

4-D22
2-D22

3-D22

3-D22
2-D22

3,000
3,000

2,2-D10@100
4Fl.
2,2-D10@140

2-D19 3-D19
4C1 4WC25

5-D19RG3 3-D19

2-D19

4C2

4C1 3-D22

RG2

3-D19

2-D13
2-D13
2-D13
2-D13
2-D13

3-D22

RFl.

2-D19

3-D22

RG1

2,2-D10@140

2-D19

3-D22
1-D22

RG1

2-D22
3-D22

3-D22

4-D22

RFl.

4-D22
2-D22

D10@200
3-D22
1-D22

Figure A.5 Steel locations at floor 1F.

77

Figure A.6 Steel locations at floor 2F.

78

Figure A.7 Steel locations at floor 2F.

79

Figure A.8 Steel locations at floor 3F.

80

Figure A.9 Steel locations at floor 3F.

81

Figure A.10 Steel locations at floor 4F.

82

Figure A.11 Steel locations at floor 4F.

83

A.2 MEMBER GEOMETRY AND REINFORCEMENT OF THE PT SPECIMEN

C

FS1
A

PG1

3,600

P0

PG2
P Wall

PC1

PC1

X

2,500

3,600

3,600

7,200
3a

3

PG1

PG2

A

2,500

Y

PC1

PS1

FS1

P Wall

P0

PG3

PC1

7,200

7,200

B

PS1

PG1

7,200

P0

14,400P014,400B

PG2
P WallFS1

7,200

P Wall
FS1

PG2

PG1

C

PC1

PC1

P0

P0

3,600
7,200

4

3

3a

(a) Foundation

4(b) 2-4 stories

Figure A.12 Floor plan of the PT specimen.

RPG1

3PC1
3PG1

3,000

3PC1
3PG1

2PC1

2PC1
2PG1

2PC1
2PG1

1PC1

1PC1

7,200

1PC1

4Fl.

4PG2

3Fl.

3PG2

2Fl.

2PG2

1Fl.

800

14,400

800

A

B

C

3

(a) 3-frame and 4-frame

Figure A.13

2PC1

2PC1

3a

2PC1
2PG3

2PG2

7,200

3PC1
3PG3

1PC1

1PC1

1PC1

3,600

3,600
1,400
4

(b) A-frame and C-frame

Elevation of the PT specimen.

84

3PC1

3PC1

P Wall

1,400

4PC1
4PG3

P Wall

1,200

450
25

4PC1

3PG2

3,600

6,700
RPG3

4PG2
P Wall

7,200

450
25

4PC1

P Wall

5,970

3PC1

3,000

4PC1
4PG1

30

4PC1
4PG1

450 2,065 2,500 2,065 450
30 30
30 30
RPG2
RPG2

6,370

1Fl.

4Fl.

12,000

2Fl.

450
25

30

RPG1

6,700

3,000

5,970
30

3Fl.

6,370

3,000

450
25 25

4PC1
4Fl.

30

3,000
1,200 3,000

12,000

3,000

4Fl.

6,700

1,200 3,000

450
25

3,600

1,400

7,200

1,400

3

3a

4

(c) B-framePC specimen

Figure A.14

Overview of the PT specimen.

85

Table A.2 List of steel reinforcement.

List of Wall

List of Column

P WallPC1
D13@100
125
4Fl.
3Fl.
2Fl.

450
TendonRebar
Hoop

8-21mm(SBPR1080/1230)
4-D19
D10@100

250

SectionSection

125

450

4Fl.
3Fl.
2Fl.
1Fl.

869

250

480

4C-1-19.3mm(SWPR7BL)

Top
BottomStirrupWeb

2 - D19
3 - D19
2-D10@100
2-D10@200

125
125

250

BottomSection

1,200
381 3812,500
D13(KSS785)@75
D13(KSS785)@75
4-D13
D13@754-D22

1Fl.

400 100

120100 280

125

400 100

120 380

400
100

Tendon480

5-D13@120
Tendon
V bar
H bar

1,200
381 3812,500

3-10-15.2mm(SWPR7B)

List of Girder
PG3

PG2

300

Tendon4C-3-15.2mm(SWPR7BL)

Top
BottomStirrupWeb

2 - D19
3 - D19
2-D10@90
2-D10@200

List of SlabTop Concrete

Section
All

300

300

Top Concrete

400 100

120100 280

Location
400 100

Section3Fl.
2Fl.

400 100

2 - D19
3 - D19
2-D10@200
2-D10@150
500
120 260 120

4C-1-15.2mm(SWPR7BL)

300

Top
Section

5-D13@120

Top
BottomStirrupWeb
Section
4Fl.

300

Tendon500
120 260 120

RFl.

500
120 260 120

SectionCenterTop Concrete869

D13(KSS785)@75
D13(KSS785)@75
4-D13
D13@75

125

PG1
End

2,250
381 3812,500

3-10-15.2mm(SWPR7B)
D13@150(double)
D13@100(double)

TendonV bar
H bar

List of Girder
Location376

D13@100

300

300

Tendon 2C-1-17.8mm(SWPR19L)
Top
BottomStirrupWeb

2 - D192 - D19
2-D10@100(KSS785)

2C-1-17.8mm(SWPR19L)
1C-17mm(SBPR930/1080)
2 - D192 - D19
2-D10@150

Depth: 130mm

Shorter direction Longer direction

CS2
CS3

Top
BottomTop
BottomTop
Bottom

D10@200

D13@200
D10@200
D10@200
D10@200
D13@200
D10@200

D10@250
D10@250
D10@250
D10@250
D13@200
D10@200

CenterTendon2c-15.2mm

86

End

2,000

Top Concrete

100
30

CS1

D10@200

300

PS1

139.1
Half Precast Panel

450 25 2,075

3PG3

Tendon1-10-15.2mm
2PC1
PCBar2-21mm
PCBar3C-21mm

3,000

170
130

5,970

200 100
300

Tendon2C-1-17.8mm

2PG2

2PG3

Mortar
1Fl.

4-D25
FG3

2-D13
2-D13
4-D25

7,200

3,600

7,200
14,400

A

3,600
7,200

C

B

3

3a

3

3a
7,200

B

4

PG1

PS1

PG1

7,200

cS2

cS1

cS1
A

PG2
cS3

cS2

PW25

PG2
cS3

6,900

(d) Floor slab

Figure A.15

Details of PT specimen.

87

1,000

PG3
cS2

7,200

4

3-D25
4-D25
4

3

(b) A-frame and C-frame(a) 3-frame and 4-frame

4-D25
3-D25

850

FG2

1,500
Bondless

FG1

1PC1
2-D22

850

Tendon2C-9-12.7mm
FG1
6-D25

1PC1

1,200

1,020 180

1PC1

1Fl.

3PC1

PW25
PCBar2-21mm
PCBar3C-21mm

2Fl.

4PG3

PW25

Mortar
1PC1
CouplerMortarTendon2C-1-17.8mm

Tendon2C-1-17.8mm

3PG2

3Fl.

250

4PC1

PW25
Coupler

2PC1
6,370

2PG1

4PG2

4Fl.

3,000

Tendon2C-3-15.2mm
2PG1

2Fl.

Tendon2C-1-17.8mm

PW25

3PC1
30

12,000

100

3PG1

2PC1

850

6,370

2PC1

3,000
1,200

3PC1
Tendon2C-3-15.2mm
3PG1

30

3,000

400

3Fl.

4PG1

3,000

3,000
30

12,000PCBar2-21mm

400

RPG3

Mortar

4PC1

3,000

5,970
500
260
120 120

Top Concrete
Tendon2C-1-19.3mm
4PG1

280
120
100

3,000

4PC1

PCBar3C-21mm

100 1,100
500600 30

RPG2

RFl.

150

4Fl.

45025

450

RPG1

150

4PC1

25

1,000

225 225
Tendon2C-1-15.2mm
RPG1

RFl.

2,500

870 760 870

400

400

225 2253PC1

25

25 4506,700

300 450 1,000

450 25(c) B-frame

225
125

225
125
100
Tendon 2-21mm
Sheath 40/44mm
Tendon 3-21mm
Sheath 40/44mm
2-D19
30

Threaded Coupler
Coupler Sheath 75/79mm L=400
Hoop. -D10@100

MortarSheath
D10
Hoop. -D10@100

PL-110x110x22
Tendon 2-21mm
Sheath 40/44mm
Tendon 3-21mm
Sheath 40/44mm

400

100

4
450

Tendon 2c-1-17.8mm(SWPR19L)
Sheath 32/36mm
PESheath
Top 2-D19
3FL

120

260
500

20 100Top 2-D19
st. -D10@90

Bottom 3-D19
PESheath
Tendon 1c-3-15.2mm(SWPR7BL)
Sheath 50/54mm
MortarFigure A.16 Details of PT beam column joint.

88

3a
2,500
1,250

1,250

869

381

381

869

Tendon 2c-10-15.2mm(SWPR7BL)
Sheath 85/92mm

Grouted coupler D22
st. -D13@125
38 262
38
262

250
175

A

600

175

D13@75-S13@75

D13@150
184

540

540

Top 2-D25
Top 3-D25

184

Plan

3a
2,500
1,250
869
D13@120 117.5
D13@150

1,250

381
381
D13@150117.5

869
D13@120

D13@150
19

Grouted coupler D22
1,030
1,030

Anchor 2-D22
Tendon 1c-10-15.2mm(SWPR7BL)
Sheath 85/92mm(Bondless)D13@75
D13@75MortarTop 6-D25
Top 2-D25

183.5

200

170

1,200

30

1FL

219

600

381

381
1,200

600

219

Web 2-D13
Bottom 5-D25
Bottom 7-D25
st. -D13@125

Figure A.17 Details of PT wall base and foundation.

89

A

250
125
125

Tendon 2c-10-15.2mm(SWPR7BL)
2-D13@150
D13@100-D13@100

MortarTop2-D19

D10@200

CS2
InsertTendon 2c-1-17.8mm(SWPR19L)
Sheath 32/36mm
Bottom 2-D19

D13@100

3a
7,200
150 150
PG3
300

B

4
Tendon 1c-10-15.2mm(SWPR7BL)
Sheath 85/92mm
D13@150-D13@100

cS2

1,000

2-D13@150

3

D10@100
D10@200
100
30
130

300

30

D10@250130 170

3FL

D10@100

PG1

PS1

PG1

7,200

cS2

cS1

cS1
A

PG2
cS3

cS2

PW25

PG2
cS3

6,900

Figure A.18

Details of PT wall floor slab interface

90

A.3 SETUP AND PLACEMENT OF THE SPECIMENS
1,500

12,400
9,400

1,500

Carrier

2,550

100

8,400

15,10010,000100

2,550Concrete stub(a) Foundation on carrier
20,400RC specimen

PT specimen(i) Longer direction of shaking table(ii) Shorter direction of shaking table(b) Elevations

West

South
East

RC specimen

PT specimen

North

PT bar(c) Fixed foundations on shaking table

Figure A.19 Set up of the specimens.

91

Figure A.20

Placement of the specimens on the shaking table.

92

93

Figure A.21 Measuring weight of the specimens.

94

Figure A.22 Weights of equipment on the buildings at the third level95

Figure A.23 Weights of equipment on the buildings at roof level.

96

Appendix B

B.1 EQUIVALENT LATERAL LOAD PROCEDURE (ASCE 7-05)
SHEAR WALL DIRECTION
Mapped MCE spectral response accelerations:Ss (g) =S1 (g) =

1.5
0.9

Fa =
Fv =
I=
R=
hi =
n=

1
1
1
6
3

At short periods
At 1 s.

Site coefficients:

Importance factor:
Response modification factor:
Story height:
Number of stories:
Design spectral response acceleration parameters:
1.5
SMS (g) =SM1 (g) =
0.9

m

4

SDS (g) = 1
SD1 (g) = 0.6

Ts (sec) 0.6
T0 (sec) 0.12

Period Calculations:
Eigenvalue analysis:

Teigen=

0.58

sec

Ct =

0.0488(for metric)

hn =
x=

12
0.75

m

ASCE 7‐05 (12.8): Ct*(hn)x

Ta (sec)=

0.315

ASCE 7‐05 (12.8): 0.1N

Ta (sec)=

0.4

Approximate period:
Table 12.8‐2:

97

9.84

ft

Cu =

1.4

Tlimit = CuTa=

0.44

T (sec)=0.440(SD1>0.6)

Seismic Response Coefficient:Cs(12.8‐
2)
0.167

Csmax
(12.8‐3)

Csmin(12.8‐5)Csmin(12.8‐6)

Weight(kN)

Vbase= Cs*W

k

0.23

0.01

0.075

1815

302.50

1

i
1
2
3
4
Total weight:

wi (kN)
441
444.5
458
471.5
1815

Cvi
0.096
0.194
0.299
0.411

Fi (kN)
29.1
58.6
90.6
124.3

Mi(kN‐m)
2745
1838
1017
373

Vbase=

302.5

Story forces:** Weight is half the full weight to find the forces per shear wall system.
Redundancy Factor ρ=

1.3

Story forces with redundancy factor
Eh=ρQEi12
34

Fi (kN)
37.8
76.2
117.7
161.6

Mi(kN‐m)
3569
2389
1323
485

Eh= V*ρ =% of weight=

393.3
21.67

kN
%

98

Mu,baseFRAME DIRECTION
Mapped MCE spectral response accelerations:Ss (g) =S1 (g) =

1.5
0.9

Fa =
Fv =
I=
R=

1
1
1
8

hi =
n=

3
4

At short periods
At 1 s.

Site coefficients:

Importance factor:
Response modification factor:
Storyheight:Number of stories:

9.84

m

Design spectral response acceleration parameters:
SMS (g) =SM1 (g) =1.5
0.9

SDS (g) = 1
SD1 (g) = 0.6

Ts (sec) 0.6
T0 (sec) 0.12

Teigen=

0.67

sec

Ct =

0.0466(for metric)

hn =
x=

12
0.9

Ta (sec)=

0.44

Ta (sec)=

0.4

Cu =
Tlimit =CuTa=
T (sec)=

1.4
0.56
0.56

Csmin
(12.8‐5)Csmin(12.8‐6)

Weight(kN)

Vbase= Cs*Wk

0.01

0.056251815

226.881.015

Period Calculations:
Eigenvalue analysis:
Approximate period:
Table 12.8‐2:

ASCE 7‐05 (12.8):
Ct*(hn)x
ASCE 7‐05 (12.8):
0.1N

Seismic response Coefficient:Cs
Csmax(12.8‐2)(12.8‐3)
0.125

0.13

99

m

(SD1>0.6)ftStory forces:
i
1
2
3
4
Total weight:

wi (kN)
441
444.5
458
471.5
1815

Cvi
0.093
0.192
0.300
0.415

Fi (kN)
21.1
43.5
68.0
94.2

Vbase=

226.9

Mi(kN‐m)
2067
1387
769
283

** Weight is half the full weight to find the forces per special moment frame.
Redundancy Factor ρ=

1.3

Story forces with redundancy factor
Eh=ρQEi12
34

Fi (kN)
27.5
56.5
88.5
122.5

Mi(kN‐m)
2687
1803
1000
367

Eh= V*ρ =% of weight=

294.9
16.25

kN
%

100

B.2

CALCULATIONS BASED ON ACI 318-08 PROVISIONS

BEAMSMaterials
Concrete:f'c =27 MPa3.9

ksi

Steel:

fy =345 MPa

50.0

ksi

mm
mm
mm
mm

118.11472.44
11.81
23.62

in
in
in
in

180000 mm2
10 mm

279.00
0.39
0.12
7.87
21.48
0.87
0.59
0.39
0.12

in2
in

5.12

in

2G1 (Frame direction)
Cross‐section
hi=Total height =
bw=
h=
Ag =
Diahoop=slab thickness
slab reinforcement

3000
12000
300
600

mm2
mm
mm
mm

Ahoop=
spacing=d=
DiaBar =

78.54
200
545.5
22

Abar=
DiaWeb =380.13 mm2
10 mm

AWeb=

78.54 mm2ts=

130 mm
D10 @ 250ln =6700 mm

in2
in
in
in
in2
in
in2

Strength check:Flexural strength
effective beam width:263.78 in

S.8.12.2 : beff=min(ln/4,bw+2*[8ts],bw+2*[1/2(clear dist. to the next web)]clear dist to the next web =101

6900 mm

271.65 inoverhanging =1400 mm

55.12

beff=
beff=
beff=

1675 mm
2380 mm
7200 mm65.94 in
93.70 in
283.46 inbeff=

1675 mm

65.94

Mn+ =

385.70 kN‐m

Mn‐ =572.20 kN‐m

Mn,center=424.60 kN‐m

Mnmax =

572.20 kN‐m

Max total length from:
1) Total length <= ln/42)Each side <= 8ts
3)
Each side <= lc/2

S21.5.2.2
mid‐spaninin

Mn+ = 385.70>

Mn‐ /2=286

OK

Mn+ = 385.70>

Mnmax /4 =

143

OK

Mn‐ = 572.20>

Mnmax /4 =

143

OK

S21.5.2.1 As,min = 3sqrt(f'c)/fy*bw*d
not less than 200bw*d/fy
As,min =614.07 mm2

0.95

in2

200bw*d/fy=

654.27 mm2

1.01

in3

1.77

in3

Try 3,6# of bars=

3

current As= 1140.40 mm2
current As = 1140.40>

check reinf. Ratioρt =ρt = 0.0070

As,min =654

OKρt ,max= 0.0250

OK

0.0070<

Shear strength

Vc=2*sqrt(f'c)*bw*dVs = Av*fy*d/s

102

Vc =
# of hoops=

141.19
2

kN

31.74

kips

Av =
Vs=

157.08147.76

mm2
kN

0.24
33.22

in3
kips

Vn=

Vc+Vs =

288.96kN

Mn,pr+ =482.13kN‐mMn,pr‐ =715.25

kN‐m

wg =

10.40

N/mm

Vn=
Vu,pr=

288.96
214

kN
kN

64.96

kips

64.96
48.01

kips
kips

214

OK

Vu,pr due to momentsΦVn = 217>

Vu =Detailing:
Transverse reinforcement
S21.5.3.1: hoops shall be provided in 2h
2h =current region length =

1200
‐

mm
mm

47.24
‐

in
in

s = min(d/4; 8db; 24dhoop; 12") =current spacing=

136.38
200

mm
mm

5.37
7.87

in
in

136

NOT OK

10.74
7.87

in
in

S21.5.3.2: max spacing in 2h:current spacing=

200>s,min =

beyond 2h:s<=d/2=current spacing=current spacing= 200

272.75200<mm
mms,min =

273

OK

3G1 (Frame direction)
Cross‐section
hi=Total height =
bw=
h=

103

3000
12000
300
600

mm
mm
mm
mm

118.11472.44
11.81
23.62

in
in
in
in

Ag =
Diahoop=slab thickness
slab reinforcement

180000 mm2
10 mm
mm2
mm
mm
mm

Ahoop=
spacing=d=
DiaBar =

78.54
200
545.5
22

Abar=
DiaWeb =380.13 mm2
10 mm

AWeb=

78.54 mm2ts=

130 mm
D10 @ 250ln =6700 mm279.00
0.39
0.12
7.87
21.48
0.87
0.59
0.39
0.12

in2
in

5.12

in

in2
in
in
in
in2
in
in2

Strength check:Flexural strength
effective beam width:263.78 in

S.8.12.2 : beff=min(ln/4,bw+2*[8ts],bw+2*[1/2(clear dist. to the next web)]clear dist to the next web =overhanging =6900 mm1400 mm

271.65 in
55.12 in

beff=
beff=
beff=

1675 mm
2380 mm
7200 mm65.94 in
93.70 in
283.46 inbeff=

1675 mm

65.94

Mn+ =

380.30 kN‐m

Mn‐ =526.90 kN‐m

Mn,center=424.60 kN‐m

Mnmax =

526.90 kN‐m

Max total length from:
1) Total length <= ln/42)Each side <= 8ts
3)
Each side <= lc/2

S21.5.2.2
mid‐spanin

Mn+ = 380.30>

Mn‐ /2=263

OK

Mn+ = 380.30>

Mnmax /4 =

132

OK

Mn‐ = 526.90>

Mnmax /4 =

132

OK

S21.5.2.1 As,min = 3sqrt(f'c)/fy*bw*d104

not less than 200bw*d/fy
As,min =614.07 mm2

0.95

in2

200bw*d/fy=

654.27 mm2

1.01

in3

1.77

in3

Try 3,5# of bars=

3

current As= 1140.40 mm2
current As = 1140.40>

check reinf. Ratioρt =ρt = 0.0070

As,min =654

OKρt ,max= 0.0250

OK

0.0070<

Shear strength
Vc=2*sqrt(f'c)*bw*dVs = Av*fy*d/s
Vn=

Vc =
# of hoops=

141.19
2

kN

31.74

kips

Av =
Vs=

157.08147.76

mm2
kN

0.24
33.22

in3
kips

Vc+Vs =

288.96

kN

64.96

kips

Mn,pr+ =

64.96
45.88

kips
kips

204

OK

Vu,pr due to moments
475.38kN‐m‐

Mn,pr =

658.63

kN‐m

wg =

10.40

N/mm

Vn=
Vu,pr=
ΦVn = 217

288.96 kN
204
kN
>

Vu =

Detailing:
Transverse reinforcement
S21.5.3.1: hoops shall be provided in 2h
2h =

105

1200

mm

47.24

in

‐

mm

‐

in

136.38
200

mm
mm

5.37
7.87

in
in

136

NOT OK

10.74
7.87

in
in

current region length =S21.5.3.2: max spacing in 2h:
s = min(d/4; 8db; 24dhoop; 12") =current spacing=current spacing= 200>s,min =

beyond 2h:s<=d/2=current spacing=272.75200

current spacing= 200

<

mm
mm
s,min =

273

OK

4G1,RG1 (Frame direction)
Cross‐section
hi=Total height =
bw=
h=
Ag =
Diahoop=slab thickness
slab reinforcement

mm
mm
mm
mm

118.11472.44
11.81
23.62

in
in
in
in

180000 mm2
10 mm

279.00
0.39
0.12
7.87
21.48
0.87
0.59
0.39
0.12

in2
in

5.12

in

3000
12000
300
600

mm2
mm
mm
mm

Ahoop=
spacing=d=
DiaBar =

78.54
200
545.5
22

Abar=
DiaWeb =380.13 mm2
10 mm

AWeb=

78.54 mm2ts=

130 mm
D10 @ 250ln =6700 mm

in2
in
in
in
in2
in
in2

Strength check:Flexural strength
effective beam width:

S.8.12.2 : beff=min(ln/4,bw+2*[8ts],bw+2*[1/2(clear dist. to the next web)]106

263.78 inclear dist to the next web =overhanging =6900 mm1400 mm

271.65 in
55.12 in

beff=
beff=
beff=

1675 mm
2380 mm
7200 mm65.94 in
93.70 in
283.46 inbeff=

1675 mm

65.94

Mn+ =

372.80 kN‐m

Mn‐ =475.40 kN‐m

Mn,center=424.60 kN‐m

Mnmax =

475.40 kN‐m

Max total length from:
1) Total length <= ln/42)Each side <= 8ts
3)
Each side <= lc/2

S21.5.2.2
mid‐spanin

Mn+ = 372.80>

Mn‐ /2=

238

OK

Mn+ = 372.80>

Mnmax /4 =

119

OK

Mn‐ = 475.40>

Mnmax /4 =

119

OK

S21.5.2.1 As,min = 3sqrt(f'c)/fy*bw*d
not less than 200bw*d/fy
As,min =614.07 mm2

0.95

in2

200bw*d/fy=

654.27 mm2

1.01

in3

1.77

in3

Try 3,4# of bars=

3

current As= 1140.40 mm2
current As = 1140.40>

check reinf. Ratioρt =ρt = 0.0070

As,min =654

OKρt ,max= 0.0250

OK

0.0070<

Shear strength
Vc=2*sqrt(f'c)*bw*d

107

Vc =
# of hoops=

141.19
2

kN

31.74

kips

Av =

157.08

mm2

0.24

in3

Vs = Av*fy*d/s
Vn=

Vs=

147.76

kN

33.22

kips

Vc+Vs =

288.96

kN

64.96

kips

Mn,pr+ =466.00kN‐mMn,pr‐ =594.25

kN‐m

wg =

10.40

N/mm
64.96
43.41

kips
kips

193

OK

Vu,pr due to moments

Vn=Vu,pr=
ΦVn = 217

288.96 kN
193
kN
>

Vu =

Detailing:
Transverse reinforcement
S21.5.3.1: hoops shall be provided in 2h
2h =current region length =

1200
‐

mm
mm

47.24
‐

in
in

s = min(d/4; 8db; 24dhoop; 12") =current spacing=

136.38
200

mm
mm

5.37
7.87

in
in

136

NOT OK

10.74
7.87

in
in

S21.5.3.2: max spacing in 2h:current spacing= 200>s,min =

beyond 2h:s<=d/2=current spacing=current spacing= 200

272.75
200
<

108

mm
mm
s,min =273

OK

CORNER COLUMNS
Materials
Concrete:f'c =27 MPa3.9

ksi

Steel:

fy =345 MPa

50.0

ksi

mm
mm
mm
mm

118.11472.44
19.69
19.69

in
in
in
in

250000 mm2
10 mm

387.50
0.39
0.12
3.94
17.54
0.87
0.59

in2
in

11.81
23.62
94.49

in
in
in

1C1 (Frame Direction) ‐‐ Corner Column
Cross‐section
hi=Total height
Hc =
Bc =
Ag=
Diahoop=
Ahoop=
spacing=d=
DiaBar =Beam(s) Connected:
2G1

3000
12000
500
500

78.54
100
445.5
22

mm2
mm
mm
mm

Abar=

380.13 mm2bw=h=hclear=R=
I=
Cd=

300 mm
600 mm
2400 mm
8
1
5.5

in2
in
in
in
in2

Strength check:Flexural strength
Column strength:Mn,coltop =

346

kN‐m

bottom

429
200

kN‐m
kN‐m

Mn,col=Mu=

ΦMn = 225

>

109

Mu =

200

OK

Beam(s) strength:Mn,beam+ =

386

kN‐m

Mn,beam‐ =

572

kN‐m

# of bars=

10

S21.6.3.1
Ast>=0.01Ag

Ast =Ast = 3801.33

3801.33

mm2

5.89

>

0.01Ag =2500.000

in2
OK

Axial Force ratio
Ptotal=P/f'cAg=

772.020.114

kN

173.57kips

Shear strength
1) Ve = 2*Mpr,col/h2) Ve =
(Mpr,beam(+)+Mpr,beam(‐))/h

Mpr,coltop =432.51 kN‐m

Mpr,colbottom =536.75 kN‐m

Ve(1) =

403.86 kN

90.80

kips

@ one axis

Ve(2) =

200.89 kN

45.16

kips

@ the other axis

Ve(2) =

298.02 kN

67.00

kips

Vu =

98.31 kN

22.10

kips

>

limit Ve/Vu =

0.5

OK

>

Agf'c/20 =337.5

NOT OK

kN

43.21

kips

kN
kN

108.56151.76kips
kips

If Ve/Vu > 0.5 & P<Agf'c/20 ‐‐> ignore Vc
current Ve/Vu = 2.04
P = 772.02
Bottom section
Vc =# of hoops=
Vs=

Vs = Av*fy*d/s
Vn= Vc+Vs
ΦVn = 506

192.184482.85675.04>

if Vs< 4(bd)sqrt(f'c) ; s < (d/2 ; 24")
if Vs< 4(bd)sqrt(f'c) ; s < (d/4 ; 12")

110

Ve =

298.02OK

check if Vs< 4(bd)sqrt(f'c)
Vs= 109s = min(d/2; 24") =current spacing=current spacing= 100>
4sqrt(f'c)bd =111.38 mm
100 mm

86
4.38
3.94

s < (d/4 ;
12)
in
in

<

s,min =

111

OK

<

8sqrt(f'c)bd =

173

OK

Vsmax = 8(bd)sqrt(f'c)
Vs= 109

Top section
Vc =# of hoops=
Vs=

Vs = Av*fy*d/s
Vn= Vc+Vs
ΦVn = 416

192.183362.14554.32

kN

43.21

kips

kN
kN

81.42
124.62

kips
kips

>

Ve =

<

4sqrt(f'c)bd =298.02

OK86

s < (d/2 ;
24)

8.77
3.94

in
in

if Vs< 4(bd)sqrt(f'c) ; s < (d/2 ; 24")
if Vs< 4(bd)sqrt(f'c) ; s < (d/4 ; 12")check if Vs< 4(bd)sqrt(f'c)
Vs= 81s = min(d/2; 24") =current spacing=current spacing= 100

222.75 mm
100 mm<s,min =

223

OK

<

8sqrt(f'c)bd =

173

OK

Vsmax = 8(bd)sqrt(f'c)
Vs= 81
Detailing:
S21.6.4# of hoops =
Ash =

111

4
314.16

mm2

0.49

in2

bc =

417

mm

16.42

in

2

Ach =

173889mm

269.53

in2

hx =
so =

226
144.8

mm
mm

8.90
5.70

in
in

400.00 mm
2400 mm

15.75
94.49

in
in

S21.6.4.1
lo <= min (member depth; 1/6*clear height; 18")lo =current lo=

spacing same everywhere‐‐>current lo= 2400>

lo,min =400

OKWithin lo:s <= min (h/4;6db; so; 6")s = min (h/4; 6db;
so; 6")
current spacing=current
spacing= 100

125 mm
100 mm

<

4.92
3.94

s,min =

in
in

125

OK

S21.6.4.4
s (1) <= Ash / (0.3 bc f'c/fy (Ag/Ach ‐ 1))
s (2) <= Ash / (0.09 bc f'c/fy)s(1) =

73.31 mm

2.89

in

(2)

106.96 mm

4.21

in

73

NOT OK

5.20
3.94

in
in

s =
current spacing= 100>s,min =

Beyond lo:
s = min (6db; 6")
current spacing=current spacing= 100

132 mm
100 mm<112

s,min =

132

OK

Drift check: (ASCE7‐05 12.12)
Δs shall be <= 0.02/ρ = 0.015
δxe
Floor
h (mm)
(mm)
22.30
4
12000

δx (mm)Δi122.650.0068

OK

3

9000

18.60

102.3

0.0108

OK

2

6000

12.70

69.85

0.0134

OK

1

3000

5.40

29.7

0.0099OK

Δtotal=113

0.0102OKINTERIOR COLUMNS
Materials
Concrete:f'c =27 MPa3.9 ksi

Steel:

fy =345 MPa

50.0 ksi

1C2 (Frame Direction) ‐‐ Interior Column
Cross‐section
hi=Total height=
Hc =
Bc =
Ag=
Diahoop=
Ahoop=
spacing=d=
DiaBar =Beam(s) Connected:
2 x 2G1

mm
mm
mm
mm

118.11472.44
19.69
19.69

in
in
in
in

250000 mm2
10 mm

387.50
0.39
0.12
3.94
17.54
0.87
0.59

in2
in

11.81
23.62
94.49

in
in
in

3000
12000
500
500

78.54
100
445.5
22

mm2
mm
mm
mm

Abar=

380.13 mm2bw=h=hclear=R=
I=
Cd=

300 mm
600 mm
2400 mm
8
1
5.5

in2
in
in
in
in2

Strength check:Flexural strength
Column strength:Mn,coltop =

486

kN‐m

bottom

486
205

kN‐m
kN‐m

Mn,col=Mu=

ΦMn = 316
Beam(s) strength:>Mn,beam+ =

114

386

Mu = 205kN‐m

OK

Mn,beam‐ =572# of bars=

10

kN‐m

S21.6.3.1
Ast>=0.01Ag

Ast = 3801.33 mm2
Ast = 3801.33>

5.89 in2

0.01Ag = 2500.000

OK

Axial Force ratio
Ptotal= 1222.22 kN
P/f'cAg= 0.181

274.78kips

Shear strength
1) Ve = 2*Mpr,col/h2) Ve =
(Mpr,beam(+)+Mpr,beam(‐))/h

Mpr,coltop =607.38 kN‐m

Mpr,colbottom =607.38 kN‐m

Ve(1) =

506.15 kN

113.79kips

Ve(2) =

498.91 kN

112.16kips

Vu =98.31 kN

22.10

kips

If Ve/Vu > 0.5 & P<Agf'c/20 ‐‐> ignore Vc
current Ve/Vu = 5.07
P = 1222.22
Vc =# of hoops=
Vc =
Vs=

Vs = Av*fy*d/s
Vn= Vc+Vs
ΦVn = 506>

Ve/Vu lim=

0.5

OK

>

Agf'c/20 =337.5

NOT OK

192.184
192.18482.85675.04

kN

43.21

kips

kN
kN
kN

43.21
108.56151.76

kips
kips
kips

>

Ve =

498.91>4sqrt(f'c)bd 86

OK

if Vs< 4(bd)sqrt(f'c) ; s < (d/2 ; 24")
if Vs< 4(bd)sqrt(f'c) ; s < (d/4 ; 12")check if Vs< 4(bd)sqrt(f'c)
Vs= 109s = min(d/2; 24") =115

111.38 mm

s < (d/4 ; 12)4.38 incurrent spacing=current spacing= 100

100 mm

3.94 in<s,min = 111

OK<

8sqrt(f'c)bd 173

OK

Vsmax = 8(bd)sqrt(f'c)
Vs= 109
Detailing:
S21.6.4# of hoops =

4

Ash =
bc =

314.16417

mm2mm

0.49 in2
16.42 inAch =173889mm2

269.53 in2

hx =
so =

240
140.1

mm
mm

9.45 in5.52 in

400.00 mm
2400 mm15.75 in94.49 inS21.6.4.1
lo <= min (member depth; 1/6*clear height; 18")lo =current lo=

spacing same everywhere‐‐>current lo= 2400>

lo,min = 400

OKWithin lo:
S21.6.4.3
s <= min (h/4; 6db; so; 6")s = min (h/4; 6db;
so; 6")
current spacing=current spacing= 100

125 mm
100 mm
<4.92 in3.94 ins,min = 125

OK

S21.6.4.4
s (1) <= Ash / (0.3 bc f'c/fy (Ag/Ach ‐ 1))
s (2) <= Ash / (0.09 bc f'c/fy)s(1) =

73.31 mm

2.89 in(2)

106.96 mm

4.21 ins =current spacing= 100>s,min = 73

NOT OK

Beyond lo:
s = min (6db; 6")
current spacing=

116

132 mm
100 mm

5.20 in3.94 incurrent spacing= 100<s,min = 132

OK

2C2 (Frame Direction) ‐‐ Interior Column
Cross‐section
hi=Total height
Hc =
Bc =
Ag=
Diahoop=
Ahoop=
spacing=d=
DiaBar =Beam(s) Connected:
2 x 3G1

mm
mm
mm
mm

118.11472.44
19.69
19.69

in
in
in
in

250000 mm2
10 mm

387.50
0.39
0.12
3.94
17.54
0.87
0.59

in2
in

11.81
23.62
94.49

in
in
in

3000
12000
500
500

78.54
100
445.5
22

mm2
mm
mm
mm

Abar=

380.13 mm2bw=h=hclear=R=
I=
Cd=

300 mm
600 mm
2400 mm
8
1
5.5

in2
in
in
in
in2

Strength check:Flexural strength
Column strength:Mn,coltop =

456

kN‐m

Mn,colbottom =
Mu=

456
187

kN‐m
kN‐m

ΦMn = 296Beam(s) strength:

>

Mu =

Mn,beam+ =

380

kN‐m

Mn,beam‐ =

527

kN‐m

# of bars=

10

187

OK

S21.6.3.1
Ast>=0.01Ag

Ast = 3801.33 mm2

117

5.89

in2

Ast = 3801.33>

0.01Ag = 2500.000

OK

Axial Force ratio
Ptotal=P/f'cAg=

919.650.136

kN

206.76

kips

Shear strength
1) Ve = 2*Mpr,col/h2) Ve =
(Mpr,beam(+)+Mpr,beam(‐))/h

Mpr,coltop =569.80 kN‐m

Mpr,colbottom =569.80 kN‐m

Ve(1) =

474.84 kN

106.75kips

Ve(2) =

472.50 kN

106.23kips

Vu =89.01 kN

20.01

kips

>

limit Ve/Vu

0.5

OK

>

Agf'c/20 =337.5

NOT OK

If Ve/Vu > 0.5 & P<Agf'c/20 ‐‐> ignore Vc
current Ve/Vu = 5.31
P = 919.65
Vc =# of hoops=
Vs=

Vs = Av*fy*d/s
Vn= Vc+Vs
ΦVn = 506

192.184482.85675.04

kN

43.21

kips

kN
kN

108.56151.76

kips
kips

>

Ve =

>

4sqrt(f'c)bd

472.50OK

86

s < (d/4 ; 12)

4.38
3.94

in
in

if Vs< 4(bd)sqrt(f'c) ; s < (d/2 ; 24")
if Vs< 4(bd)sqrt(f'c) ; s < (d/4 ; 12")check if Vs< 4(bd)sqrt(f'c)
Vs= 109s = min(d/2; 24") =current spacing=current spacing= 100

111.38 mm
100 mm<Vsmax = 8(bd)sqrt(f'c)

118s,min =111

OK

Vs= 109<8sqrt(f'c)bd

173

OK

Detailing:
S21.6.4# of hoops =

4

Ash =
bc =

314.16
407

mm2
mm

0.49
16.02

in2
in

Ach =

165649mm2

256.76

in2

hx =
so =

163
165.8

mm
mm

6.42
6.53

in
in

400.00 mm
2400 mm

15.75
94.49

in
in

S21.6.4.1
lo <= min (member depth; 1/6*clear height; 18")lo =current lo=

spacing same everywhere‐‐>current lo= 2400>

lo,min =400

OKWithin lo:s <= min (h/4; 6db; so; 6")s = min (h/4; 6db; so; 6")
current spacing=current spacing= 100

125 mm
100 mm<4.92
3.94
s,min =

in
in

125

OK

S21.6.4.4
s (1) <= Ash / (0.3 bc f'c/fy (Ag/Ach ‐ 1))
s (2) <= Ash / (0.09 bc f'c/fy)s(1) =

64.56 mm2.54

ins(2) =

109.59 mm

4.31

in

current spacing= 100>s,min =65

NOT OK

Beyond lo:
s = min (6db; 6")
current spacing=119

132 mm
100 mm

5.20
3.94

in
in

current spacing= 100<s,min =

132

OK

3C2 (Frame Direction) ‐‐ Interior Column
Cross‐section
hi=Total height
Hc =
Bc =
Ag=
Diahoop=
Ahoop=
spacing=d=
DiaBar =Beam(s) Connected:
2 x 4G1

mm
mm
mm
mm

118.11472.44
19.69
19.69

in
in
in
in

250000 mm2
10 mm

387.50
0.39
0.12
3.94
17.54
0.87
0.59

in2
in

11.81
23.62
94.49

in
in
in

3000
12000
500
500

78.54
100
445.5
22

mm2
mm
mm
mm

Abar=

380.13 mm2bw=h=hclear=R=
I=
Cd=

300 mm
600 mm
2400 mm
8
1
5.5

in2
in
in
in
in2

Strength check:Flexural strength
Column strength:Mn,coltop =

442

kN‐m

bottom

442
153

kN‐m
kN‐m

Mn,col=Mu=

ΦMn = 287
Beam(s) strength:>

Mu = 153

Mn,beam+ =

373

kN‐m

Mn,beam‐ =

475

kN‐m

120

OK

S21.6.3.1
Ast>=0.01Ag# of bars=

10

Ast = 3801.33 mm2
Ast = 3801.33>

5.89 in2

0.01Ag = 2500.000

OK

Axial Force ratio
Ptotal=P/f'cAg=

620.640.092

kN

139.53kips

Shear strength
1) Ve = 2*Mpr,col/h2) Ve =
(Mpr,beam(+)+Mpr,beam(‐))/h

Mpr,coltop =552.85 kN‐m

Mpr,colbottom =552.85 kN‐m

Ve(1) =

460.71 kN

103.58kips

Ve(2) =

441.77 kN

99.32

kips

Vu =

70.07 kN

15.75

kips

If Ve/Vu > 0.5 & P<Agf'c/20 ‐‐> ignore Vc
current Ve/Vu = 6.30
P = 620.64

Vs = Av*fy*d/s>

limit Ve/Vu

0.5

OK

>

Agf'c/20 =337.5

NOT OK

Vc =# of hoops=

192.18
2

kN

43.21

kips

Vs=

241.43433.61

kN
kN

54.28
97.48

kips
kips

Vn= Vc+Vs
ΦVn = 325<

Ve =441.77<4sqrt(f'c)bd 86

NOT OK

if Vs< 4(bd)sqrt(f'c) ; s < (d/2 ; 24")
if Vs< 4(bd)sqrt(f'c) ; s < (d/4 ; 12")check if Vs< 4(bd)sqrt(f'c)
Vs= 54s = min(d/2; 24") =121

222.75 mm

s < (d/2 ; 24)
8.77 incurrent spacing=current spacing= 100

100 mm

3.94 in<s,min = 223

OK<

8sqrt(f'c)bd 173

OK

Vsmax = 8(bd)sqrt(f'c)
Vs= 54

Detailing:
S21.6.4# of hoops =

2

Ash =
bc =

157.08409

mm2
mm

0.24 in2
16.10 inAch =167281mm2

259.29 in2

hx =
so =

210
150.1

mm
mm

8.27 in5.91 in

400.00 mm
2400 mm15.75 in94.49 inS21.6.4.1
lo <= min (member depth; 1/6*clear height; 18")lo =current lo=

spacing same everywhere‐‐>current lo= 2400>

lo,min = 400

OKWithin lo:s <= min (h/4; 6db; so; 6")s = min (h/4; 6db; so; 6")
current spacing=current spacing= 100

125 mm
100 mm
<4.92 in3.94 ins,min = 125

OK

S21.6.4.4
s (1) <= Ash / (0.3 bc f'c/fy (Ag/Ach ‐ 1))
s (2) <= Ash / (0.09 bc f'c/fy)s(1) =

33.08 mm1.30 in(2)

54.53 mm2.15 ins =current spacing= 100>

122s,min = 33

NOT OKBeyond lo:
s = min (6db; 6")
current spacing=current spacing= 100

132 mm
100 mm
<5.20 in3.94 ins,min = 132

Beam Column Joint ‐ G1‐C2‐G1 ‐ frame direction (case 1)
Materials
Concrete:f'c =3.9 ksi

Steel:fy =50 ksi

Bslab =
B=
d=

66 in
11.81 in22.10 inn=3As,1 =

0.60 in2

As =1.8 in2

a = As fy / (0.85 f'c B) =0.41 in

Cross‐section

Nominal Moment Capacity of Beams ‐ G1

Mn+

Mn ‐

#7
bars

Mn+ =

1970.49 in‐kip

Mn+ =164.21 ft‐kipn=4

As,1 =

0.6 in2As =2.4 in2

a = As fy / (0.85 f'c B) =3.07 in

#7
bars

Mn‐ =

2468.09 in‐kip

Mn‐ =205.67 ft‐kip

Mn,pr+ =205.26 ft‐kip‐

Mn,pr =

123

257.09 ft‐kip

OKInterior Connection G1‐C2‐G1
hcolumn =19.68 inbcol =

19.68 inbw =x=

11.81 in3.94 inbeff =

19.68 indb =0.875 in

Asb,1 =

1.8 in2

Asb,2 =

2.4 in2f'c =3900 psi

γV =

12

#7
Long beam bars: bars

(beams frame into three faces of a column but
the beam width is less than 3/4 of the column width)

Mpr,b1 =257.09 ft‐kip

Mpr,b2 =205.26 ft‐kip

hclear =8.86 in

MC1 = MC2 = MC = (Mpr,b1 +Mpr,b2)/2 =231.18 ft‐kip

VC1 = MC1 / (hclear/2) =52.19 kip

Vu,joint = 1.25 fy Asb,1 + 1.25 fy
Asb,2 ‐VC1 =210.31 kip

Aj =387.30 in2

θV =

0.85

Aj =

246.71 ksi

Joint Shear Demand

θV Vn = θV γV (fc')

0.5

Vu,joint

θV Vn<

Joint Detailing Requirements

124

=>

OK

bw =
=>

11.81

<

3/4 bcol =14.76

Required transferse reinforcement = 100% Ash
3 #3
bars

Column ‐ C2

longitudial column
bars:

Ash =

0.33 in2

Ach =216.97 in2

Ag =387.30 in2

bc =14.56 inhx =7.905 indb,col =0.875 inso =6.03 ins < Ash / (0.3 bc f'c/fy (Ag/Ach ‐1)) =1.23 ins = min(b/4; 6db; so; 6") =4.92 insA =5.52 in#7

Actual spacing in the structure:
Joint Anchorage Requirements

Mn+ / Mn‐ =0.80> 0.5,
OK

Beam longitudial reinforcement should be extended to the far face of the confined column andanchored in tension.
ldh = fy db / (65 (f'c)0.5) =ldh,req=

10.78> 8 db =

6 "

7.0

or 6"

ldh,act =14 "OKBeam Column Joint ‐ G1 ‐ C1 ‐ frame direction (case 2)
Materials
Concrete:

f'c =

125

3.9 ksiSteel:fy =50 ksi

Bslab =
B=
d=

66 in
11.81 in22.10 inn=3As,1 =

0.60 in2

As =1.80 in2

a = As fy / (0.85 f'c B) =0.41 in

Cross‐section

Nominal Moment Capacity of Beams ‐ G1
#7
bars

Mn+

Mn

Mn+ =

1970.49 in‐kip

Mn+ =164.21 ft‐kipn=4

As,1 =

0.60 in2As =2.4 in2

a = As fy / (0.85 f'c B) =3.07 in

#7
bars

‐

Mn‐ =

2468.09 in‐kip

Mn‐ =205.67 ft‐kip

Mn,pr+ =205.26 ft‐kip

Mn,pr‐ =257.09 ft‐kip

hcolumn =19.68 inbcol =

19.68 inbw =x=

11.81 in3.94 inbeff =

19.68 indb =0.875 inAsb,2 =

2.40 in2

f'c =3900 psiγV =12Exterior Connection G1 ‐ C1

#7Long beam bars: bars

126

(beams frame into two faces of a column)

Mpr,b1 =257.09 ft‐kip

hclear =8.86 in

MC1 = MC2 = MC = (Mpr,b)/2 =128.55 ft‐kip

VC1 = MC1 / (hclear/2) =29.02 kip

Vu,joint = 1.25 fy Asb,2 ‐VC1 =120.98 kip

Aj =387.30 in2

θV =

0.85

Aj =

246.71 ksi

Joint Shear Demand

θV Vn = θV γV (fc')

0.5

Vu,joint

<

θV Vn

=>

OK

11.81

<

3/4 bcol =14.76

Joint Detailing Requirementsbw ==>Required transferse reinforcement = 100% Ash

Column ‐ C1

longitudial column
bars:

3 #3
bars

Ash =

0.33 in2

Ach =216.97 in2

Ag =387.30 in2

bc =14.56 inhx =7.905 indb,col =0.875 inso =6.03 ins < Ash / (0.3 bc f'c/fy (Ag/Ach ‐ 1)) =1.23 ins = min(b/4; 6db; so; 6") =4.92 insA =5.52 in#7

Actual spacing in the structure:127

Joint Anchorage Requirements
Mn+ / Mn‐ =0.80> 0.5,
OK‐ beam longitudial reinforcement should be extended to the far face of the confined column
and anchored in tension.
ldh = fy db / (65 (f'c)0.5) =ldh,req=

10.78> 8 db =10.78 "

7.0

or 6"

ldh,act =

14 "
OK

128

WALLS
Materials
Concrete:f'c =27 MPa3.9

ksi

Steel:

fy =345 MPa

50.0

ksi

mm
mm
mm
mm

118.11472.44
98.43
9.84

in
in
in
in

625000 mm2
10 mm

968.75
0.39
0.12
3.94
4.92
7.87
15.75

in2
in

Cross‐section
hi=
hw=
Lw =
tw=
Acv=
Diahoop=
Ahoop=
hoop spacing=
web transverse spacing (AXIS A)=
web transverse spacing (AXIS C)=
boundary width =# of bars in the boundary=Diabar=

3000
12000
2500
250

78.54
100
125
200
400
6
19

mm2
mm
mm
mm
mm
mm

Abar=
R=
I=
Cd=

283.53 mm
6
1
5

Mn =
Mu=

2884
3569

2

in2
in
in
in
in

0.75
0.44

in

3569

NOT OK

in2

Strength check:Flexural strengthΦMn = 2595.4

kN‐m
kN‐m

<

Shear strength
AXIS Aαc=# of hoops=(transverse reinforcement ratio)
ρt=ρt= 0.0050

22
0.0050>

Vn= 1622.74 kN
393
Vu=
kN‐m

129

Mu =

ρmin= 0.0025

OK

364.83 kips

ΦVn = 1217

>

Vu =

393

OK

ρmin= 0.0025

OK

AXIS Cαc=# of hoops=(transverse reinforcement ratio)
ρt=ρt= 0.0031

22
0.0031>

Vn= 1216.42 kN
393
Vu=
kN‐m
ΦVn = 912>

273.48 kips
Vu =

393

OK

Axial Force ratio
Ptotal=P/f'cAg=

284.86
0.017

kN

64

kips

δxe=
δ u=
δu/hw=

28.46

1.12

142.320.0119

mm
mm

5.60

in
in

lw/600(δu/hw)=
c=351.33243.50

mm
mm

13.83
9.59

in
in

351

BE NOT
NEEDED

Detailing:
Need for special boundary elements:
At design‐based earthquake DBE(elastic displacement)
δu/hw shall not be less than 0.007‐‐>
check ifc>=lw/600(δu/hw)(from BIAX)

c= 244
At maximum considered earthquake MCE(elastic displacement)

climit =δxe=δ u=
δu/hw=

42.69
213.470.0178

mm
mm

1.68
8.40

in
in

lw/600(δu/hw)=c=234.22243.50

mm
mm

9.22
9.59

in
in

δu/hw shall not be less than 0.007‐‐>
check if<c>=lw/600(δu/hw)

c= 244>130lw/600(δu/hw)
234
=

BE
NEEDED‐‐>

if boundary elements are needed, length of BE:
c'=larger of {c‐0.1lw,c/2}

c'=current BE length =121.75400mm
mm‐‐> if not needed, satisfy 21.9.6.5
21.9.6.5(a): if ρ> 400/fy ; satisfy 21.6.4.2 and 21.9.6.4(a); s<8"400/fy=ρ= 0.0170‐‐

0.0080>400/fy= 0.0080‐> 21.6.4.2 and 21.9.6.4(a) ; s<821.9.6.4(a) : c'=larger of {c‐0.1lw,c/2}
c'=current BE length =121.75
400

mm
mm

183.0

mm

‐‐ 21.6.4.2: hx<14"
1st floor
current hx =

hx =
183.0

upper floors
current hx =

<
hx=

275.0

275.0

7.20
hxlimit =

<

355.6
10.83

mm

in
OK
in

hxlimit =

355.6

OK

8 in=
8db=

203.2
152

OK
OK

‐‐ check spacing: s < 8db ;8"s=current s = 100
current s = 100

100 mm<<Hoop reinforcement Ash:in x‐dir:# of hoops=
bc=

2
163

Ach =

51345

mm

6.42
79.58

Ag =

100000mm2

155.00 in2

131

mm
2

in
in2

157

mm2

0.24

min Ash =

363
<

mm2

0.56 in2
min Ash = 362.65 NOT OK

min Ash =

115
>

mm2

0.18 in2
min Ash = 114.81
OK# of hoops=
bc=

3
315

mm

12.40
0.37

current Ash=in2

(eq.21‐4) Ash >= 0.3 s bc f'c /fy (Ag/Ach ‐ 1)
current Ash= 157(eq. 21‐5) Ash >= 0.09 s bc f'c/fy
current Ash= 157
in y‐dir:current Ash=

2

in
in2

236

mm

min Ash =

701
<

mm2

1.09 in2
min Ash = 700.82 NOT OK

min Ash =

115
>

mm2

0.18 in2
min Ash = 114.81
OK(eq.21‐4) Ash >= 0.3 s bc f'c /fy (Ag/Ach ‐ 1)
current Ash= 236(eq. 21‐5) Ash >= 0.09 s bc f'c/fy
current Ash= 236

Drift check: (ASCE7‐05 12.12)
Δs shall be <= 0.02/ρ = 0.015
Floor

h (mm)
4
3
2
1

12000
9000
6000
3000

δxe
(mm)
28.46
18.49
9.44
2.68
Δtotal=δx (mm)Δi142.32
92.45
47.19
13.42

0.01660.01510.01130.00450.0119OK

NOTE: Member capacities are calculated based on SD345 strength for all reinforcement.

132

NOT OK
OK
OK
OK

Appendix C

C.1 CONSTRUCTION PROCESS

Figure C.1

Construction of RC specimen versus PT specimen.

133

Figure C.2

Construction of RC specimen.

134

Figure C.3

Construction of PT specimen (column).

135

Figure C.4

Construction of PT specimen (beam and slab).

136

Figure C.5 Construction of PT specimen (walls).

137

Figure C.6

Construction of PT specimen (walls).

138

Appendix D

D.1 INSTRUMENTATION
3

2

4

7,200

C

1

7,200

7,200

7,200

A

B

7,200

7,200

B

(i) Plan(i) Plan

2,870

3,000

130

3Fl.

2,870

3,000

130

2Fl.

1Fl.

7,200

7,200

7,200
B

A

C

B

(ii) Elevation(a)(ii) Elevation

RC specimen(b)

Displacement transducer

Figure D.1

Accelerometer

Measurements.

139

7,200

PC specimen

140

141

142

143

144

145

146

147

148

149

150

151

152

153

154

155

156

157

158

159

160

161

162

163

164

165

166

167

168

169

170

171

172

173

174

175

176

177

178

179

180

181

182

183

184

185

186

187

188

189

190

191

192

193

194

195

196

197

198

199

200

201

202

203

204

205

206

207

208

209

210

211

212

213

214

215

216

217

218

219

220

221

222

223

224

225

226

Appendix E

PSA (cm/s2)

E.1 PSEUDO ACCELERATION SPECTRA OF THE GROUND MOTIONS

Figure E.1

Acceleration spectra for JMA-Kobe ground motion (x-direction).

227

E.2

Acceleration spectra for JMA-Kobe ground motion (y-direction).

E.3

Acceleration spectra for Takatori ground motion (x-direction).

228

Figure E.4

Acceleration spectra for Takatori ground motion (y-direction).

E.2 PSEUDO VELOCITY SPECTRA OF THE GROUND MOTIONS

Figure E.5

Pseudo velocity spectra for JMA-Kobe ground motion (x-direction).

229

Figure E.6

Pseudo velocity spectra for JMA-Kobe ground motion (y-direction)

Figure E.7

Pseudo velocity spectra for Takatori ground motion (x-direction)

230

Figure E.8

Pseudo velocity spectra for Takatori ground motion (y-direction)

E.3 DISPLACEMENT SPECTRA OF THE GROUND MOTIONS

Figure E.9

Displacement spectra for the Kobe ground motion (x-direction).

231

Figure E.10

Displacement spectra for the Kobe ground motion (y-direction)

Figure E.11

Displacement spectra for the Takatori ground motion (x-direction)

232

Figure E.12

Displacement spectra for the Takatori ground motion (y-direction)

233

234

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Toward Earthquake-Resistant Design of Concentrically Braced Steel Structures. Patxi Uriz and Stephen A. Mahin.
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Integrated Probabilistic Performance-Based Evaluation of Benchmark Reinforced Concrete Bridges. Kevin R.
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PEER 2006/07

Analytical Modeling of Reinforced Concrete Walls for Predicting Flexural and Coupled–ShearFlexural Responses. Kutay Orakcal, Leonardo M. Massone, and John W. Wallace. October 2006.

PEER 2006/06

Nonlinear Analysis of a Soil-Drilled Pier System under Static and Dynamic Axial Loading. Gang Wang andNicholas Sitar. November 2006.

PEER 2006/05

Advanced Seismic Assessment Guidelines. Paolo Bazzurro, C. Allin Cornell, Charles Menun, Maziar Motahari,and Nicolas Luco. September 2006.

PEER 2006/04

Probabilistic Seismic Evaluation of Reinforced Concrete Structural Components and Systems. Tae Hyung Lee
and Khalid M. Mosalam. August 2006.

PEER 2006/03

Performance of Lifelines Subjected to Lateral Spreading. Scott A. Ashford and Teerawut Juirnarongrit. July 2006.

PEER 2006/02

Pacific Earthquake Engineering Research Center Highway Demonstration Project. Anne Kiremidjian, James
Moore, Yue Yue Fan, Nesrin Basoz, Ozgur Yazali, and Meredith Williams. April 2006.

PEER 2006/01

Bracing Berkeley. A Guide to Seismic Safety on the UC Berkeley Campus. Mary C. Comerio, Stephen Tobriner,and Ariane Fehrenkamp. January 2006.

PEER 2005/16

Seismic Response and Reliability of Electrical Substation Equipment and Systems. Junho Song, Armen Der
Kiureghian, and Jerome L. Sackman. April 2006.

PEER 2005/15

CPT-Based Probabilistic Assessment of Seismic Soil Liquefaction Initiation. R. E. S. Moss, R. B. Seed, R. E.
Kayen, J. P. Stewart, and A. Der Kiureghian. April 2006.

PEER 2005/14

Workshop on Modeling of Nonlinear Cyclic Load-Deformation Behavior of Shallow Foundations. Bruce L. Kutter,
Geoffrey Martin, Tara Hutchinson, Chad Harden, Sivapalan Gajan, and Justin Phalen. March 2006.

PEER 2005/13

Stochastic Characterization and Decision Bases under Time-Dependent Aftershock Risk in Performance-Based
Earthquake Engineering. Gee Liek Yeo and C. Allin Cornell. July 2005.

PEER 2005/12

PEER Testbed Study on a Laboratory Building: Exercising Seismic Performance Assessment. Mary C. Comerio,
editor. November 2005.

PEER 2005/11

Van Nuys Hotel Building Testbed Report: Exercising Seismic Performance Assessment. Helmut Krawinkler,
editor. October 2005.

PEER 2005/10

First NEES/E-Defense Workshop on Collapse Simulation of Reinforced Concrete Building Structures. September
2005.

PEER 2005/09

Test Applications of Advanced Seismic Assessment Guidelines. Joe Maffei, Karl Telleen, Danya Mohr, William
Holmes, and Yuki Nakayama. August 2006.

PEER 2005/08

Damage Accumulation in Lightly Confined Reinforced Concrete Bridge Columns. R. Tyler Ranf, Jared M. Nelson,
Zach Price, Marc O. Eberhard, and John F. Stanton. April 2006.

PEER 2005/07

Experimental and Analytical Studies on the Seismic Response of Freestanding and Anchored Laboratory
Equipment. Dimitrios Konstantinidis and Nicos Makris. January 2005.

PEER 2005/06

Global Collapse of Frame Structures under Seismic Excitations. Luis F. Ibarra and Helmut Krawinkler. September
2005.

PEER 2005//05

Performance Characterization of Bench- and Shelf-Mounted Equipment. Samit Ray Chaudhuri and Tara C.
Hutchinson. May 2006.

PEER 2005/04

Numerical Modeling of the Nonlinear Cyclic Response of Shallow Foundations. Chad Harden, Tara Hutchinson,
Geoffrey R. Martin, and Bruce L. Kutter. August 2005.

PEER 2005/03

A Taxonomy of Building Components for Performance-Based Earthquake Engineering.
September 2005.

PEER 2005/02

Fragility Basis for California Highway Overpass Bridge Seismic Decision Making. Kevin R. Mackie and Božidar
Stojadinović. June 2005.

PEER 2005/01

Empirical Characterization of Site Conditions on Strong Ground Motion. Jonathan P. Stewart, Yoojoong Choi,and Robert W. Graves. June 2005.

PEER 2004/09

Electrical Substation Equipment Interaction: Experimental Rigid Conductor Studies. Christopher Stearns andAndré Filiatrault. February 2005.

PEER 2004/08

Seismic Qualification and Fragility Testing of Line Break 550-kV Disconnect Switches. Shakhzod M. Takhirov,
Gregory L. Fenves, and Eric Fujisaki. January 2005.

PEER 2004/07

Ground Motions for Earthquake Simulator Qualification of Electrical Substation Equipment. Shakhzod M.
Takhirov, Gregory L. Fenves, Eric Fujisaki, and Don Clyde. January 2005.

PEER 2004/06

Performance-Based Regulation and Regulatory Regimes. Peter J. May and Chris Koski. September 2004.

PEER 2004/05

Performance-Based Seismic Design Concepts and Implementation: Proceedings of an International Workshop.
Peter Fajfar and Helmut Krawinkler, editors. September 2004.

PEER 2004/04

Seismic Performance of an Instrumented Tilt-up Wall Building. James C. Anderson and Vitelmo V. Bertero. July
2004.

PEER 2004/03

Evaluation and Application of Concrete Tilt-up Assessment Methodologies. Timothy Graf and James O. Malley.
October 2004.

PEER 2004/02

Analytical Investigations of New Methods for Reducing Residual Displacements of Reinforced Concrete Bridge
Columns. Junichi Sakai and Stephen A. Mahin. August 2004.

PEER 2004/01

Seismic Performance of Masonry Buildings and Design Implications. Kerri Anne Taeko Tokoro, James C.
Anderson, and Vitelmo V. Bertero. February 2004.

PEER 2003/18

Performance Models for Flexural Damage in Reinforced Concrete Columns. Michael Berry and Marc Eberhard.
August 2003.

PEER 2003/17

Predicting Earthquake Damage in Older Reinforced Concrete Beam-Column Joints. Catherine Pagni and Laura
Lowes. October 2004.

PEER 2003/16

Seismic Demands for Performance-Based Design of Bridges. Kevin Mackie and Božidar Stojadinović. August
2003.

PEER 2003/15

Seismic Demands for Nondeteriorating Frame Structures and Their Dependence on Ground Motions. Ricardo
Antonio Medina and Helmut Krawinkler. May 2004.

Keith A. Porter.

PEER 2003/14

Finite Element Reliability and Sensitivity Methods for Performance-Based Earthquake Engineering. Terje
Haukaas and Armen Der Kiureghian. April 2004.

PEER 2003/13

Effects of Connection Hysteretic Degradation on the Seismic Behavior of Steel Moment-Resisting Frames. Janise
E. Rodgers and Stephen A. Mahin. March 2004.

PEER 2003/12

Implementation Manual for the Seismic Protection of Laboratory Contents: Format and Case Studies. William T.
Holmes and Mary C. Comerio. October 2003.

PEER 2003/11

Fifth U.S.-Japan Workshop on Performance-Based Earthquake Engineering Methodology for Reinforced
Concrete Building Structures. February 2004.

PEER 2003/10

A Beam-Column Joint Model for Simulating the Earthquake Response of Reinforced Concrete Frames. Laura N.
Lowes, Nilanjan Mitra, and Arash Altoontash. February 2004.

PEER 2003/09

Sequencing Repairs after an Earthquake: An Economic Approach. Marco Casari and Simon J. Wilkie. April 2004.

PEER 2003/08

A Technical Framework for Probability-Based Demand and Capacity Factor Design (DCFD) Seismic Formats.
Fatemeh Jalayer and C. Allin Cornell. November 2003.

PEER 2003/07

Uncertainty Specification and Propagation for Loss Estimation Using FOSM Methods. Jack W. Baker and C. Allin
Cornell. September 2003.

PEER 2003/06

Performance of Circular Reinforced Concrete Bridge Columns under Bidirectional Earthquake Loading. Mahmoud
M. Hachem, Stephen A. Mahin, and Jack P. Moehle. February 2003.

PEER 2003/05

Response Assessment for Building-Specific Loss Estimation. Eduardo Miranda and Shahram Taghavi.
September 2003.

PEER 2003/04

Experimental Assessment of Columns with Short Lap Splices Subjected to Cyclic Loads. Murat Melek, John W.
Wallace, and Joel Conte. April 2003.

PEER 2003/03

Probabilistic Response Assessment for Building-Specific Loss Estimation. Eduardo Miranda and Hesameddin
Aslani. September 2003.

PEER 2003/02

Software Framework for Collaborative Development of Nonlinear Dynamic Analysis Program. Jun Peng andKincho H. Law. September 2003.

PEER 2003/01

Shake Table Tests and Analytical Studies on the Gravity Load Collapse of Reinforced Concrete Frames. Kenneth
John Elwood and Jack P. Moehle. November 2003.

PEER 2002/24

Performance of Beam to Column Bridge Joints Subjected to a Large Velocity Pulse. Natalie Gibson, André
Filiatrault, and Scott A. Ashford. April 2002.

PEER 2002/23

Effects of Large Velocity Pulses on Reinforced Concrete Bridge Columns. Greg L. Orozco and Scott A. Ashford.
April 2002.

PEER 2002/22

Characterization of Large Velocity Pulses for Laboratory Testing. Kenneth E. Cox and Scott A. Ashford. April
2002.

PEER 2002/21

Fourth U.S.-Japan Workshop on Performance-Based Earthquake Engineering Methodology for Reinforced
Concrete Building Structures. December 2002.

PEER 2002/20

Barriers to Adoption and Implementation of PBEE Innovations. Peter J. May. August 2002.

PEER 2002/19

Economic-Engineered Integrated Models for Earthquakes: Socioeconomic Impacts. Peter Gordon, James E.
Moore II, and Harry W. Richardson. July 2002.

PEER 2002/18

Assessment of Reinforced Concrete Building Exterior Joints with Substandard Details. Chris P. Pantelides, Jon
Hansen, Justin Nadauld, and Lawrence D. Reaveley. May 2002.

PEER 2002/17

Structural Characterization and Seismic Response Analysis of a Highway Overcrossing Equipped with
Elastomeric Bearings and Fluid Dampers: A Case Study. Nicos Makris and Jian Zhang. November 2002.

PEER 2002/16

Estimation of Uncertainty in Geotechnical Properties for Performance-Based Earthquake Engineering. Allen L.
Jones, Steven L. Kramer, and Pedro Arduino. December 2002.

PEER 2002/15

Seismic Behavior of Bridge Columns Subjected to Various Loading Patterns. Asadollah Esmaeily-Gh. and Yan
Xiao. December 2002.

PEER 2002/14

Inelastic Seismic Response of Extended Pile Shaft Supported Bridge Structures. T.C. Hutchinson, R.W.
Boulanger, Y.H. Chai, and I.M. Idriss. December 2002.

PEER 2002/13

Probabilistic Models and Fragility Estimates for Bridge Components and Systems. Paolo Gardoni, Armen Der
Kiureghian, and Khalid M. Mosalam. June 2002.

PEER 2002/12

Effects of Fault Dip and Slip Rake on Near-Source Ground Motions: Why Chi-Chi Was a Relatively Mild M7.6
Earthquake. Brad T. Aagaard, John F. Hall, and Thomas H. Heaton. December 2002.

PEER 2002/11

Analytical and Experimental Study of Fiber-Reinforced Strip Isolators. James M. Kelly and Shakhzod M. Takhirov.
September 2002.

PEER 2002/10

Centrifuge Modeling of Settlement and Lateral Spreading with Comparisons to Numerical Analyses. Sivapalan
Gajan and Bruce L. Kutter. January 2003.

PEER 2002/09

Documentation and Analysis of Field Case Histories of Seismic Compression during the 1994 Northridge,
California, Earthquake. Jonathan P. Stewart, Patrick M. Smith, Daniel H. Whang, and Jonathan D. Bray. October
2002.

PEER 2002/08

Component Testing, Stability Analysis and Characterization of Buckling-Restrained Unbonded Braces .
Cameron Black, Nicos Makris, and Ian Aiken. September 2002.

PEER 2002/07

Seismic Performance of Pile-Wharf Connections. Charles W. Roeder, Robert Graff, Jennifer Soderstrom, and Jun
Han Yoo. December 2001.

PEER 2002/06

The Use of Benefit-Cost Analysis for Evaluation of Performance-Based Earthquake Engineering Decisions.
Richard O. Zerbe and Anthony Falit-Baiamonte. September 2001.

PEER 2002/05

Guidelines, Specifications, and Seismic Performance Characterization of Nonstructural Building Components andEquipment. André Filiatrault, Constantin Christopoulos, and Christopher Stearns. September 2001.

PEER 2002/04

Consortium of Organizations for Strong-Motion Observation Systems and the Pacific Earthquake Engineering
Research Center Lifelines Program: Invited Workshop on Archiving and Web Dissemination of GeotechnicalData, 4–5 October 2001. September 2002.

PEER 2002/03

Investigation of Sensitivity of Building Loss Estimates to Major Uncertain Variables for the Van Nuys Testbed.
Keith A. Porter, James L. Beck, and Rustem V. Shaikhutdinov. August 2002.

PEER 2002/02

The Third U.S.-Japan Workshop on Performance-Based Earthquake Engineering Methodology for Reinforced
Concrete Building Structures. July 2002.

PEER 2002/01

Nonstructural Loss Estimation: The UC Berkeley Case Study. Mary C. Comerio and John C. Stallmeyer.
December 2001.

PEER 2001/16

Statistics of SDF-System Estimate of Roof Displacement for Pushover Analysis of Buildings. Anil K. Chopra,
Rakesh K. Goel, and Chatpan Chintanapakdee. December 2001.

PEER 2001/15

Damage to Bridges during the 2001 Nisqually Earthquake. R. Tyler Ranf, Marc O. Eberhard, and Michael P.
Berry. November 2001.

PEER 2001/14

Rocking Response of Equipment Anchored to a Base Foundation. Nicos Makris and Cameron J. Black.
September 2001.

PEER 2001/13

Modeling Soil Liquefaction Hazards for Performance-Based Earthquake Engineering. Steven L. Kramer andAhmed-W. Elgamal. February 2001.

PEER 2001/12

Development of Geotechnical Capabilities in OpenSees. Boris Jeremić. September 2001.

PEER 2001/11

Analytical and Experimental Study of Fiber-Reinforced Elastomeric Isolators. James M. Kelly and Shakhzod M.
Takhirov. September 2001.

PEER 2001/10

Amplification Factors for Spectral Acceleration in Active Regions. Jonathan P. Stewart, Andrew H. Liu, Yoojoong
Choi, and Mehmet B. Baturay. December 2001.

PEER 2001/09

Ground Motion Evaluation Procedures for Performance-Based Design. Jonathan P. Stewart, Shyh-Jeng Chiou,
Jonathan D. Bray, Robert W. Graves, Paul G. Somerville, and Norman A. Abrahamson. September 2001.

PEER 2001/08

Experimental and Computational Evaluation of Reinforced Concrete Bridge Beam-Column Connections for
Seismic Performance. Clay J. Naito, Jack P. Moehle, and Khalid M. Mosalam. November 2001.

PEER 2001/07

The Rocking Spectrum and the Shortcomings of Design Guidelines. Nicos Makris and Dimitrios Konstantinidis.
August 2001.

PEER 2001/06

Development of an Electrical Substation Equipment Performance Database for Evaluation of Equipment
Fragilities. Thalia Agnanos. April 1999.

PEER 2001/05

Stiffness Analysis of Fiber-Reinforced Elastomeric Isolators. Hsiang-Chuan Tsai and James M. Kelly. May 2001.

PEER 2001/04

Organizational and Societal Considerations for Performance-Based Earthquake Engineering. Peter J. May. April
2001.

TM

PEER 2001/03

A Modal Pushover Analysis Procedure to Estimate Seismic Demands for Buildings: Theory and Preliminary
Evaluation. Anil K. Chopra and Rakesh K. Goel. January 2001.

PEER 2001/02

Seismic Response Analysis of Highway Overcrossings Including Soil-Structure Interaction. Jian Zhang and Nicos
Makris. March 2001.

PEER 2001/01

Experimental Study of Large Seismic Steel Beam-to-Column Connections. Egor P. Popov and Shakhzod M.
Takhirov. November 2000.

PEER 2000/10

The Second U.S.-Japan Workshop on Performance-Based Earthquake Engineering Methodology for Reinforced
Concrete Building Structures. March 2000.

PEER 2000/09

Structural Engineering Reconnaissance of the August 17, 1999 Earthquake: Kocaeli (Izmit), Turkey. Halil Sezen,
Kenneth J. Elwood, Andrew S. Whittaker, Khalid Mosalam, John J. Wallace, and John F. Stanton. December
2000.

PEER 2000/08

Behavior of Reinforced Concrete Bridge Columns Having Varying Aspect Ratios and Varying Lengths ofConfinement. Anthony J. Calderone, Dawn E. Lehman, and Jack P. Moehle. January 2001.

PEER 2000/07

Cover-Plate and Flange-Plate Reinforced Steel Moment-Resisting Connections. Taejin Kim, Andrew S. Whittaker,
Amir S. Gilani, Vitelmo V. Bertero, and Shakhzod M. Takhirov. September 2000.

PEER 2000/06

Seismic Evaluation and Analysis of 230-kV Disconnect Switches. Amir S. J. Gilani, Andrew S. Whittaker, Gregory
L. Fenves, Chun-Hao Chen, Henry Ho, and Eric Fujisaki. July 2000.

PEER 2000/05

Performance-Based Evaluation of Exterior Reinforced Concrete Building Joints for Seismic Excitation. Chandra
Clyde, Chris P. Pantelides, and Lawrence D. Reaveley. July 2000.

PEER 2000/04

An Evaluation of Seismic Energy Demand: An Attenuation Approach. Chung-Che Chou and Chia-Ming Uang. July
1999.

PEER 2000/03

Framing Earthquake Retrofitting Decisions: The Case of Hillside Homes in Los Angeles. Detlof von Winterfeldt,
Nels Roselund, and Alicia Kitsuse. March 2000.

PEER 2000/02

U.S.-Japan Workshop on the Effects of Near-Field Earthquake Shaking. Andrew Whittaker, ed. July 2000.

PEER 2000/01

Further Studies on Seismic Interaction in Interconnected Electrical Substation Equipment. Armen Der Kiureghian,
Kee-Jeung Hong, and Jerome L. Sackman. November 1999.

PEER 1999/14

Seismic Evaluation and Retrofit of 230-kV Porcelain Transformer Bushings. Amir S. Gilani, Andrew S. Whittaker,
Gregory L. Fenves, and Eric Fujisaki. December 1999.

PEER 1999/13

Building Vulnerability Studies: Modeling and Evaluation of Tilt-up and Steel Reinforced Concrete Buildings. John
W. Wallace, Jonathan P. Stewart, and Andrew S. Whittaker, editors. December 1999.

PEER 1999/12

Rehabilitation of Nonductile RC Frame Building Using Encasement Plates and Energy-Dissipating Devices.
Mehrdad Sasani, Vitelmo V. Bertero, James C. Anderson. December 1999.

PEER 1999/11

Performance Evaluation Database for Concrete Bridge Components and Systems under Simulated Seismic
Loads. Yael D. Hose and Frieder Seible. November 1999.

PEER 1999/10

U.S.-Japan Workshop on Performance-Based Earthquake Engineering Methodology for Reinforced Concrete
Building Structures. December 1999.

PEER 1999/09

Performance Improvement of Long Period Building Structures Subjected to Severe Pulse-Type Ground Motions.
James C. Anderson, Vitelmo V. Bertero, and Raul Bertero. October 1999.

PEER 1999/08

Envelopes for Seismic Response Vectors. Charles Menun and Armen Der Kiureghian. July 1999.

PEER 1999/07

Documentation of Strengths and Weaknesses of Current Computer Analysis Methods for Seismic Performance ofReinforced Concrete Members. William F. Cofer. November 1999.

PEER 1999/06

Rocking Response and Overturning of Anchored Equipment under Seismic Excitations. Nicos Makris and Jian
Zhang. November 1999.

PEER 1999/05

Seismic Evaluation of 550 kV Porcelain Transformer Bushings. Amir S. Gilani, Andrew S. Whittaker, Gregory L.
Fenves, and Eric Fujisaki. October 1999.

PEER 1999/04

Adoption and Enforcement of Earthquake Risk-Reduction Measures. Peter J. May, Raymond J. Burby, T. Jens
Feeley, and Robert Wood.

PEER 1999/03

Task 3 Characterization of Site Response General Site Categories. Adrian Rodriguez-Marek, Jonathan D. Bray,and Norman Abrahamson. February 1999.

PEER 1999/02

Capacity-Demand-Diagram Methods for Estimating Seismic Deformation of Inelastic Structures: SDF Systems.
Anil K. Chopra and Rakesh Goel. April 1999.

PEER 1999/01

Interaction in Interconnected Electrical Substation Equipment Subjected to Earthquake Ground Motions. Armen
Der Kiureghian, Jerome L. Sackman, and Kee-Jeung Hong. February 1999.

PEER 1998/08

Behavior and Failure Analysis of a Multiple-Frame Highway Bridge in the 1994 Northridge Earthquake. Gregory L.
Fenves and Michael Ellery. December 1998.

PEER 1998/07

Empirical Evaluation of Inertial Soil-Structure Interaction Effects. Jonathan P. Stewart, Raymond B. Seed, andGregory L. Fenves. November 1998.

PEER 1998/06

Effect of Damping Mechanisms on the Response of Seismic Isolated Structures. Nicos Makris and Shih-Po
Chang. November 1998.

PEER 1998/05

Rocking Response and Overturning of Equipment under Horizontal Pulse-Type Motions. Nicos Makris andYiannis Roussos. October 1998.

PEER 1998/04

Pacific Earthquake Engineering Research Invitational Workshop Proceedings, May 14–15, 1998: Defining theLinks between Planning, Policy Analysis, Economics and Earthquake Engineering. Mary Comerio and Peter
Gordon. September 1998.

PEER 1998/03

Repair/Upgrade Procedures for Welded Beam to Column Connections. James C. Anderson and Xiaojing Duan.
May 1998.

PEER 1998/02

Seismic Evaluation of 196 kV Porcelain Transformer Bushings. Amir S. Gilani, Juan W. Chavez, Gregory L.
Fenves, and Andrew S. Whittaker. May 1998.

PEER 1998/01

Seismic Performance of Well-Confined Concrete Bridge Columns. Dawn E. Lehman and Jack P. Moehle.
December 2000.

ONLINE REPORTS
The following PEER reports are available by Internet only at http://peer.berkeley.edu/publications/peer_reports.html
PEER 2011/104 Design and Instrumentation of the 2010 E-Defense Four-Story Reinforced Concrete and Post-Tensioned
Concrete Buildings. Takuya Nagae, Kenichi Tahara, Taizo Matsumori, Hitoshi Shiohara, Toshimi Kabeyasawa,
Susumu Kono, Minehiro Nishiyama (Japanese Research Team) and John Wallace, Wassim Ghannoum, Jack
Moehle, Richard Sause, Wesley Keller, Zeynep Tuna (U.S. Research Team). June 2011.
PEER 2011/103 In-Situ Monitoring of the Force Output of Fluid Dampers: Experimental Investigation. Dimitrios Konstantinidis,
James M. Kelly, and Nicos Makris. April 2011.
PEER 2011/102 Ground-motion prediction equations 1964 - 2010. John Douglas. April 2011.
PEER 2011/101 Report of the Eighth Planning Meeting of NEES/E-Defense Collaborative Research on Earthquake Engineering.
Convened by the Hyogo Earthquake Engineering Research Center (NIED), NEES Consortium, Inc. February
2011.
PEER 2010/111 Modeling and Acceptance Criteria for Seismic Design and Analysis of Tall Buildings. Task 7 Report for the Tall
Buildings Initiative - Published jointly by the Applied Technology Council. October 2010.
PEER 2010/110 Seismic Performance Assessment and Probabilistic Repair Cost Analysis of Precast Concrete Cladding Systems
for Multistory Buildlings. Jeffrey P. Hunt and Božidar Stojadinovic. November 2010.
PEER 2010/109 Report of the Seventh Joint Planning Meeting of NEES/E-Defense Collaboration on Earthquake Engineering.
Held at the E-Defense, Miki, and Shin-Kobe, Japan, September 18–19, 2009. August 2010.
PEER 2010/108 Probabilistic Tsunami Hazard in California. Hong Kie Thio, Paul Somerville, and Jascha Polet, preparers. October
2010.
PEER 2010/107 Performance and Reliability of Exposed Column Base Plate Connections for Steel Moment-Resisting Frames.
Ady Aviram, Božidar Stojadinovic, and Armen Der Kiureghian. August 2010.
PEER 2010/106 Verification of Probabilistic Seismic Hazard Analysis Computer Programs. Patricia Thomas, Ivan Wong, andNorman Abrahamson. May 2010.
PEER 2010/105 Structural Engineering Reconnaissance of the April 6, 2009, Abruzzo, Italy, Earthquake, and Lessons Learned. M.
Selim Günay and Khalid M. Mosalam. April 2010.
PEER 2010/104 Simulating the Inelastic Seismic Behavior of Steel Braced Frames, Including the Effects of Low-Cycle Fatigue.
Yuli Huang and Stephen A. Mahin. April 2010.
PEER 2010/103 Post-Earthquake Traffic Capacity of Modern Bridges in California. Vesna Terzic and Božidar Stojadinović. March2010.
PEER 2010/102 Analysis of Cumulative Absolute Velocity (CAV) and JMA Instrumental Seismic Intensity (IJMA) Using the PEER–NGA Strong Motion Database. Kenneth W. Campbell and Yousef Bozorgnia. February 2010.
PEER 2010/101 Rocking Response of Bridges on Shallow Foundations. Jose A. Ugalde, Bruce L. Kutter, and Boris Jeremic. April
2010.
PEER 2009/109 Simulation and Performance-Based Earthquake Engineering Assessment of Self-Centering Post-Tensioned
Concrete Bridge Systems. Won K. Lee and Sarah L. Billington. December 2009.
PEER 2009/108 PEER Lifelines Geotechnical Virtual Data Center. J. Carl Stepp, Daniel J. Ponti, Loren L. Turner, Jennifer N. Swift,
Sean Devlin, Yang Zhu, Jean Benoit, and John Bobbitt. September 2009.
PEER 2009/107 Experimental and Computational Evaluation of Current and Innovative In-Span Hinge Details in Reinforced
Concrete Box-Girder Bridges: Part 2: Post-Test Analysis and Design Recommendations. Matias A. Hube andKhalid M. Mosalam. December 2009.
PEER 2009/106 Shear Strength Models of Exterior Beam-Column Joints without Transverse Reinforcement. Sangjoon Park andKhalid M. Mosalam. November 2009.
PEER 2009/105 Reduced Uncertainty of Ground Motion Prediction Equations through Bayesian Variance Analysis. Robb Eric S.
Moss. November 2009.
PEER 2009/104 Advanced Implementation of Hybrid Simulation. Andreas H. Schellenberg, Stephen A. Mahin, Gregory L. Fenves.
November 2009.
PEER 2009/103 Performance Evaluation of Innovative Steel Braced Frames. T. Y. Yang, Jack P. Moehle, and Božidar
Stojadinovic. August 2009.

PEER 2009/102 Reinvestigation of Liquefaction and Nonliquefaction Case Histories from the 1976 Tangshan Earthquake. Robb
Eric Moss, Robert E. Kayen, Liyuan Tong, Songyu Liu, Guojun Cai, and Jiaer Wu. August 2009.
PEER 2009/101 Report of the First Joint Planning Meeting for the Second Phase of NEES/E-Defense Collaborative Research on
Earthquake Engineering. Stephen A. Mahin et al. July 2009.
PEER 2008/104 Experimental and Analytical Study of the Seismic Performance of Retaining Structures. Linda Al Atik and Nicholas
Sitar. January 2009.
PEER 2008/103 Experimental and Computational Evaluation of Current and Innovative In-Span Hinge Details in Reinforced
Concrete Box-Girder Bridges. Part 1: Experimental Findings and Pre-Test Analysis. Matias A. Hube and Khalid M.
Mosalam. January 2009.
PEER 2008/102 Modeling of Unreinforced Masonry Infill Walls Considering In-Plane and Out-of-Plane Interaction. Stephen
Kadysiewski and Khalid M. Mosalam. January 2009.
PEER 2008/101 Seismic Performance Objectives for Tall Buildings. William T. Holmes, Charles Kircher, William Petak, and Nabih
Youssef. August 2008.
PEER 2007/101 Generalized Hybrid Simulation Framework for Structural Systems Subjected to Seismic Loading. Tarek Elkhoraibi
and Khalid M. Mosalam. July 2007.
PEER 2007/100 Seismic Evaluation of Reinforced Concrete Buildings Including Effects of Masonry Infill Walls. Alidad Hashemi
and Khalid M. Mosalam. July 2007.

The Pacific Earthquake Engineering Research Center (PEER) is a multi-institutional research andeducation center with headquarters at the University of California, Berkeley. Investigators from over 20
universities, several consulting companies, and researchers at various state and federal government
agencies contribute to research programs focused on performance-based earthquake engineering.
These research programs aim to identify and reduce the risks from major earthquakes to life safety andto the economy by including research in a wide variety of disciplines including structural and geotechnicalengineering, geology/seismology, lifelines, transportation, architecture, economics, risk management, andpublic policy.
PEER is supported by federal, state, local, and regional agencies, together with industry partners.

PEER reports can be ordered at http://peer.berkeley.edu/publications/peer_reports.html or by contacting
Pacific Earthquake Engineering Research Center
University of California, Berkeley
325 Davis Hall, mail code 1792
Berkeley, CA 94720-1792
Tel: 510-642-3437
Fax: 510-642-1655
Email: peer_editor@berkeley.edu
ISSN 1547-0587X