Bridge Engineering Institute Conference in 2023 (BEI-2023)
Rome, Italy
July 17 to 20, 2023

Edited by
Yail J. Kim
Isamu Yoshitake
Vanissorn Vimonsatit
Xuhui He
Yongcheng Ji
Jun Wang

ISBN: 978-1-7340386-1-3
The Bridge Engineering Institute
An International Technical Society
www.beibridge.org

Table of Contents
The Bridge Engineering Institute, An International Technical Society…………...………………9Conference Organization………………………………………………………………………...10
Plenary Speakers…………………………………………..…………………………………….11
Plenary Speeches
A Review of 30 Years of Experience in Field Testing of Bridges………………………….....16
Issam Harik
Performance-Based Electrochemical Analysis of an Innovative Magnesium Phosphate Cement
Coating to Mitigate Mild Steel Bar Corrosion…………………………………………..…......19
John J. Myers, Fan Zhang, and Hongyan Ma
Diagnostic Load Testing of a Bridge Before and After Rehabilitation………………………….24
Jiayi Ding, Nicolas Begasse, Adriana Toro, Francisco De Caso, and Antonio Nanni
Current State of Expressways and New Technologies for Bridge Deck Maintenance in Japan..29
Takashi Yamane
Materials and Performance
Quality of Lamination Structure Made by 3D Printing under Different Lamination Conditions.35
Yuji Yamada, Tomoko Fukuyama, Yunmi Kim, Hiroharu Kamada, and Dhruva N. Katpady
Bond Strength Evaluation of Grout Materials for Closure Pour of Precast Concrete Bridge Deck
Panels……………………………………………….…..............................................................39
Zhifu Yang
Experimental Investigation of the Effect of Debonding on PSC Girders…..……………............44
Shambhavi Dube and Durgesh C. Rai
Investigation of Nitrogen Oxide Removal of Cementitious Materials Containing TiO2……….49
Byoungsun Park, Suji Woo, Young Cheol Choi, and Sung-Won Yoo
The Diagnosis Method for Neutralization of Concrete Using Microorganisms…………………53
China Kuratomi, Atsushi Teramoto, So Fujiyoshi, and Fumito Maruyama
Research on Fatigue Properties of Basalt Fiber Cement Stabilized Macadam……………..........58
Wei Li, Peifeng Cheng, Wenmei Zhao, and Zhanming Zhang
Wind, Vibration, and Aerodynamics
Analysis of Dynamic Characteristics of Semi-Submersible Deep-Water Floating Foundation
under Wind-Wave Coupling Actions…….………………………………………………………64
Honggang Xu, Haiquan Jing, and Xuhui He

1

Field Measurement Study on Wind Characteristics at the Bridge Site of Plateau Deep-Cut
Canyon…………………………………………………………………………………………...69
Xinghui Kang, Yunfeng Zou, Xuhui He, and Longan Li
Dynamical Analysis of Vehicle-Bridge Interaction System under Ice and Wind Loads………..74
Tianyu Wu, Wenliang Qiu, Guowen Yao, and Zengwei Guo
Safety Analysis of Wind Barriers on the Bridge under the 400 km/h High-Speed Trains Loadingof Slipstream……………………………………………………………………………………..82
Dianyi Guo, Yunfeng Zou, Xuhui He, and Haobo Liang
Fatigue Analysis of Anchored Concrete Structures of Catenary Induced by Train Wind Load
Considering Different Bolt Pre-Tightening…….………………………………………………..87
Zhuole Shu, Pengru Deng, Hanfeng Wang, and Xuhui He
Influence of Spacing on Aerodynamic Interference of Parallel Rail-Cum-Road Bridges with
Twin Separated Parallel Girder…………………………………………………………………..92
Jing He, Yunfeng Zou, Xuhui He, and Lulu Liu
Monitoring
A Trust-Based Approach for Resilient UAV Networks Monitoring Bridges……….…………...97
Francesco Buccafurri and Francesca Scoleri
Evaluation Method for Grouting Condition in Post-tensioned PC Bridges Considering Directive
Propagation Properties of Elastic Waves in the Wide-Range Ultrasonic Testing (WUT)……101
Takanori Kinoshita, Yoshino Sako, Kuniharu Fukushima, and Isamu Yoshitake
Detection of Defects in Prestressed Concrete Bridge Box Girders Using GPR Image
Reconstruction…………………………………….……………………………………………105
Wael Zatar, Hien Nghiem, and Hai Nguyen
Prestressing Force Distribution in Post-Tensioned I-Girder Bridges Measured Using Smart
Strands……………………………………….…………………………………………………110
Se-Jin Jeon, Sang-Hyun Kim, Sung Yong Park, and Sung Tae Kim
Detection of Cable Section Loss in Cable-stayed Bridge with Electromagnetic Sensor……….115
Imjong Kwahk, Changbin Joh, Ji-Young Choi, Kwang-Yeun Park, and Joo-Hyung Lee
Evaluation of Salt Resistance of Three-Component Cement Concrete by UnsteadyElectrophoresis Test………………………..………………………………………………….119
Kohei Kanaoka and Tatsuya Nukushina
Field Testing
Grating on Bridge Deck for Improving Flutter Performance: A Case Study of Cable-Suspended
Bridge………………………..…………………………………………………………………125
Zhen Wang, Xuhui He, Haiquan Jing, and Lulu Liu2

Development of Elevated Road and Rail Projects in Singapore…………………………..……129
Hooi Leng Phua
Structural Behavior of a 40+ Year Old PSC-I Concrete Bridge According to its Internal Tendon
Loss in the Central Part………………………………………………………………………...133
Yeonghun Seong and Hyunjoong Kim
The Use of Engineered Cementitious Composite (ECC) Deck Overlay at Curry Street Bridge,
Windsor, Ontario, Canada : Lessons Learned and Challenges…………………..…...……..…134
Don Gardonio, Steve Volpatti, Emad Booya, and Philip Loh
Transporting Heavy Weight Cargo, Diagnosis Philosophy, and a Case Study………………..143
Daniel Elmaleh and Yehonatan Pestes
Development of Pre-Tension Method for On-Site Production of Prestressed Concrete Girders
with Span Length between 20 m to 50 m………………………….……………………….….149
Dong-Woo Seo, Sangki Park, Ki-Tae Park, Hyun-Ock Jang, and Yeon-Woo Shin
Environmental Effects
Temperature-Induced Variation in Geometric Characteristics of Existing Cracks in Double-Box
Girders of Concrete Cable-Stayed Bridge………………………………………..……………155
Xujia Liu, Youliang Ding, Hanwei Zhao, Chunfeng Wan, and Fangfang Geng
Utilizing Pressurized Injection of Lithium Nitrite in Treating Alkali-Silica Reaction and Salt
Damage in Concrete Structures………………………………………………………………160
Shohei Notsu, Tatsuya Kitada, Son Van Nguyen, and Kazunori Era
Fatigue of Steel Anchor Box on Cable-Stayed Bridges Considering Thermal and Vortex-Induced
Vibration under Random Traffic Flows………………………………………………………164
Zhiwen Zhu, Shuang Yan, Federico Accornero, and Yichun Deng
Mechanism of Control Method for the Alkali-Silica Reaction and Delayed Ettringite Formation
Expansion Using Artificial Lightweight Aggregate…………………….……………………170
Atsushi Teramoto and Moka Tanabe
Repair and FRP Composites
Electromechanical Response of Smart Repair Material with Carbon Black Based on Rapid
Setting Cement…………………………………………………………………..……………..175
Tae Uk Kim, Min Kyoung Kim, Lai Thanh Tu, and Dong Joo Kim
Bending Tests of GFRP Sandwich Panel Using a Rigid Polyethylene Core………..…………179
Hiroki Iwata, Sumitaka Inoue, and Masahide Matsumura
Analysis of Flexural-Bond Behavior of Beams Reinforced with CFRP………………..……183
Dongkyu Lim and Myoungsung Choi

3

Evaluation of Tensile Performance of CFRP Cable Anchorage System Using Non-Corrosive
Steel Material……………………………………………………………………………….…187
Seung-Hyeon Hwang, Ji-Young Kim, Jin-Young Yoon, Tae-Kyun Kim, and Woo-Tai Jung
Failure Mode and Adhesive Strength of Single Lap Joint of GFRP Laminated Plates withDifferent Fiver Volumes Bonded by a Low Elastic Adhesive………………………….……191
Masahide Matsumura, Sumitaka Inoue, Hiroki Iwata, and Yuto Anan
Ultra-High Performance Concrete
Titanium Alloy Reinforced Ultra-High Performance Concrete (TARUHPC): A Next Generationof Advanced Materials for Civil Infrastructure………………………………………….……196
Mahesh Acharya, Jared Cantrell, Luis Bedrinana, and Mustafa Mashal
Design of Ultra-High Performance Concrete Bridges………………………………………..201
Benjamin Graybeal and Rafic Helou
Axial Compression Performance of UHPC-Filled Square Duplex Stainless Steel Tube Short
Columns…………………………………………..…………………………………………….205
Hongyuan Tang, Xiaowei Hu, Jinjun Liu, and Duan Hong
Seismic Retrofit of Precast and Cast-in-Place Piers Using Ultra High-Performance Concrete(UHPC)…………………………………………………………………………………………210
Kathryn Hogarth, Manish Acharya, Arya Ebrahimpour, and Mustafa Mashal
Railway Bridges
Numerical Investigation of Aerodynamic Characteristics of High-Speed Railway Train-Bridge
System under Tornado-Like Winds……………………………………………………………215
Xuhui He and Simin Zou
Effects of Splitter Plates on the VIV Performances of Twin Separated Parallel Decks for a RailCum-Road Bridge……………………………………………………….…………………….219
Lulu Liu, Xuhui He, Yunfeng Zou, and Zhen Wang
Aerodynamics of a High-speed Train Crossing the Wake of a Bridge Tower in Different Yaw
Angles ………………………………………………………………………………………….224
Zuyu Xie, Huan Li, Xuhui He, and Zhirui Huang
Concrete Structure Damage Analysis of the Anchoring End of Wind Barriers of High-Speed
Railway Bridge under Wind Load of Train…………………………………………………..225
Haobo Liang, Yunfeng Zou, Xuhui He, and Dianyi Guo
Parametric Study on the Aerodynamic Characteristics of Wind-Guide Barriers for Train-Bridge
System…………………………………………………………………………………………..230
Shuo Jiang, Yunfeng Zou, Xuhui He, and Dianyi Guo

4

Earthquake and Dynamics
Experimental Evaluation on Seismic Behavior of Precast Concrete Walls with Developed
Vertical Joints for Enhanced Shear Strength………………………………………………....236
Kyo Young Moon, Sung Jig Kim, and Chunho Jang
Hybrid Testing of CFST Columns in Rigid-Frame Bridges Subjected to Horizontal and Vertical
Ground Motions………………………………………..…………………………………...…240
Riadh Al-Mahaidi, Javad Hashemi, and Ali Al-Attraqchi
Seismic Performance of Precast Segmental Bridge Columns with Resettable Sliding Joints: Joint
Responses and Simulation …………………………………………………………………..…244
Feng Liang, Yingqi Liu, and Francis T.K. Au
Effects of Vertical Ground Motions on the Response of Long-Span Prestressed Bridges…….249
Hsiao-Hui Hung, Chang-Wei Huang, and Yi-Hui Lu
Seismic Performance Assessment of Wall Piers Supported by Pile Foundation………………254
Shin-Tai Song and Wen-Li Huang
Experimental Techniques
Flexural Behavior of 3D-Printed Beam Members……………………..……………………….259
In-Hwan Yang, Quang-The Bui, Jihun Park, Jungwoo Lee, and Changbin Joh
ELSS Joint: Development of Semi-Rigid Joint for Joining Precast Prestressed Concrete Slabs inJapan…………………..……………………………………………………………………….264
Mohammad Emran Nasery, Tatsuhiko Mimoto, Nobuaki Sakurai
Horizontal-Vertical Ratio for Concrete Pumping Pipe …………….………………………...268
Ki-Yeol Kim, Young-Jin Kim, and Myoung-Sung Choi
Over Torque Test Method to Assess the Concrete Strength…………………………..……...273
Kenji Tada, Tatsuki Hatakeyama, and Tetsuya Ohmura
Application of Automatic Welding Robot in Integrated Construction of Bridge Pile
Foundation…………………………………………………………………………………….277
Liangjun Hu, Xiaoli Sun, and Jun Yang
Structural Behavior of Precast Concrete Beam-Column Joints…………………………….......282
Jaehyung Heo, DongIk Shin, Yeon Je Choi, Dong-Hwan Kim, Hyeong-Gook Kim, Kil-Hee Kim,
and Jung-Yoon Lee
Diaphragm Wall Mono-Shaft as an Economical Time Saving Bridge Deep Foundation…….286
Yoni Pestes and Daniel Elmaleh

5Performance Comparison of Nonlinear Dampers for Suppressing Vortex-Induced Vibration of
Flexible Bluff Structures………………..…………………………………………………….292
Jingwei Zhang, Xiaojun Wei, and Xuhui He
Control the Torsional Flutter of a Rectangular Plate with Flexible Membrane at its Leading Edge………………………………………………………………………………………………......296
Ziqiang Zhang, Hanfeng Wang, Zhiwei Liu, and Chi Zhang
Modeling and Advanced Analysis
Unseating of Bridge Decks- Codal Provisions and Fragility Analysis…………………….…302
Shambhavi Dube, Nirav Thakkar, and Durgesh C. Rai
Finite Element Modeling of Compressive Membrane Action in Transversely Prestressed Deck
Slabs…………………………………………………………………………………………….309
Amir Sana
Modelling of Bend Corners of FRP Reinforcement for Concrete Structures Considering Material
Anisotropy……..……………………………………………………………………………313
Yuanzhang Yang and Weijian Zhao
Water Pressure Variation on Porous Calcium-Silicate-Hydrates Using Molecular Dynamics…………………………………………………………………………………………………318
Shota Takinami, Ryo Yoshida, and Ryo Kobayashi
Real Imperfections of Longitudinally Stiffened Plates Compared to the Proposed Imperfections
in EN 1993-1-5……………………………………………………………………………….323
N. Thomas, A. Weinhuber, M. Mensinger, C. Holst, and J. Ndogmo
Structural Damage Prediction with Multi-Level Neural Networks Using Modal Group Response……………………………………………………………………………………..……………327
Yunfeng Zou, Xuandong Lu, Xuhui He, and Chenzhi Cai
Use of Steel Orthotropic Box Girders for Light Weight Long Span Bridges- Analysis and Designof Pedestrian Bridge in Dubai…..……………………………………………………………..331
Tharun John Joseph, Anirudh Desai, Koyya Satyanarayana, Bogdan Barbulescu, and Kannaw
Shrivatsa
Analytical Study of Fracture Behavior of Post-Tension Anchorage Zone Using FEM………..336
Kyun-Tae Lee, Jun-Mo Yang, and Jin-Kook Kim
Development of Shrinkage and Creep Prediction Model for SCC (B4-TW-SCC) and
Implementation in Bridge Design Software……………………………………………………340
Wen-Cheng Liao and Jenn-Chuan Chern

6

A Data-Driven Method for Predicting the Temperature Response of Composite Bridges Based on
Deep Learning Technology……………………………………………………………………..348
Yanjia Wang, Dong Yang, and Francis T.K. Au
Topology Optimization of Concrete Bridges Prestressed with Shape Memory Alloys………..353
Minsoo Sung and Bassem Andrawes
A Mesoscale Probabilistic Fatigue Crack Initiation Model for Steel Bridges Based on the
Modified Fine and Bhat Theory…………….………………………………….……………....358
Fei Jiang, You-Liang Ding, Kang Yang, and Xiao-Nan Zhang
Tension Estimation Method for Linked Suspenders Based on Identified Mode-Shapes…….363
Chien-Chou Chen, Wen-Hwa Wu, and Yen-Chih Peng
Research on Crack Damage Model and Size Effect of Flexible Fiber Reinforced Recycled
Concrete…......………………………………………………………………………………….368
Yongcheng Ji, Yanmin Jia, Dayang Wang, Yanwei Jia, and Zhiyang Pei
Finite Element Analysis Model of Prestressed Concrete Girder through Nonlinear Material
Properties Correction of Strand……………………………………………………………….373
Do-Yeon Kim, Young-Ji Park, Jin-Woong Choi, Young-Jin Kim, Il-Young Jang, Seong-Kyum
Kim
Corrosion and Durability
Visualization of the Effect of Steel Corrosion Cracks on Shear Deformation inside Mortar Using
X-ray CT……….……………………………………………………………………………….378
Takayuki Fumoto, Koshiro Maekawa, Shintaro Kikuchi, and Yuta Yamada
A Change in the Failure Mode and Decrease in the Strength of RC Slabs due to the Progress of
Frost Damage………………………………………………………………………………..…382
Hiroshi Hayashida
Experimental Performance of Lightweight Aggregate Concrete due to Freezing and Thawing………………………………………………………………………………………………….387
Sangwoo Kim, Sardorbek Rustamov, and Jinsup Kim
Rust Evaluation of Weathering Steel Bridges by Optical Spectra in the Visible and Near-Infrared
Regions………...……………………………………………………………………………..391
Ryuichi Inoue, Rina Hasuike, Hirokazu Furuki, and Toshihiko Aso
Effect of Measures for Inhibiting Galvanic Corrosion in an Aluminum Alloy Bridge and Guard
Railings……………………………………………………………………………………….395
Yoshito Itoh and Tatsuya Kawabata
Study of Water Absorption Properties in Actual Structure Concrete Based on Pore Structure.400
N’da Yacoub Bouadou, Peter Kuira Macharia, Lai Lai Mon, and Ryo Yoshida

7Performance Evaluation
Evaluation of the National Bridge Inventory (NBI) to Predict Bridge Deterioration in Missouri,
USA………………………………………….……………………………………………….405
Glenn Washer, John J. Myers, and Mohammed Hammed
Assessment of Live Load Carrying Capacity of Prestressed Concrete Bridge: a Case Study.…409
Shin-Tai Song and Yiching Lin
Strategic Asset Management of Locally Owned Bridges………………………………….....413
Mi G. Chorzepa
Platform and Bigdata-Driven Bridge Smart Maintenance Techniques………………….……417
Ki-Tae Park, Kyusan Jeong, Jaehwan Kim, Kun-Soo Kim, Byeongcheol Kim, and Dongwoo Seo

8The Bridge Engineering Institute, An International Technical Society
Executive Committee

Yail Jimmy Kim
President
University of Colorado DenverUSA

Isamu Yoshitake
Vice-President
Yamaguchi University
Japan

Vanissorn Vimonsatit
Director
Macquarie University
Australia

Diversity, Equity, and Inclusion (DEI) Committee

Su Taylor
Queen’s University Belfast
United Kingdom

Catherine ArmwoodGordon
Tennes. State Univ.
USA

Monique Head
University of Delaware
USA

Xuhui He
Director
Central South University
China

Eva Lantsoght
Delft Univ. of Technology
Netherlands

Jun Wang
Univ. of Colorado
Denver USA

International Advisory Committee and Secretary

Riadh Al-Mahaidi
Swinburne University ofTechnology, Australia

Brahim Benmokrane
University of Sherbrooke,
CanadaSteve C.S.Cai
Louisiana State
University, USA

Nien-Yin Chang
University of Colorado
Denver, USA

Mark F. Green
Queen’s University,
CanadaIssam Harik
University of Kentucky,
USA

Venkatesh Kodur
Michigan State
University, USA

Urs Meier
EMPA,
Switzerland

Hiroshi Mutsuyoshi
Saitama University,
Japan

John Myers
Missouri Universityof S. and T., USA

Antonio Nanni
University of Miami,
USA

Steve Nolan,
Florida Department ofTransportation, USA

Saiid Saiidi
University of Nevada
Reno, USA

Jim Shiau
University of Southern
Queensland, Australia

Johan L. Silfwerbrand
KTH Royal Institute ofTechnology, Sweden

Jongsung Sim
Hanyang University,
Korea

Ertugrul Tacioglu
University of California
Los Angeles, USA

Dan Tobias
Illinois Department ofTransportation, USA

Mark Williams
Walter P. Moore
USA

Yongcheng Ji
Northeast Forestry
University, China

9

Conference Organization
Conference Chair

Yail Jimmy Kim
University of Colorado Denver (USA)

Organizing Committee

Isamu Yoshitake (Chair)
Yamaguchi University (Japan)
Vanissorn Vimonsatit
Curtin University (Australia)
Xuhui He (China)

Central South University (China)

International Scientific Committee
Riadh S Al-Mahaidi (Australia)
Toshihiko Aso (Japan)
Brahim Benmokrane (Canada)
Steve Cai (USA)
NY Chang (USA)Mark Green (Canada)Issam Harik (USA)
Tatsumasa Kaita (Japan)
Andreas Kappos (UAE)
Venkatesh Kodur (USA)Urs Meier (Switzerland)
Hiroshi Mutsuyoshi (Japan)
John Myers (USA)
Hideaki Nakamura (Japan)

Antonio Nanni (USA)
Sriram Narasimhan (USA)Steven Nolan (USA)Saiid Saiidi (USA)Xianming Shi (USA)Jim Shiau (Australia)
Johan L Silfwerbrand (Sweden)
Jongsung Sim (Korea)
Ertugrul Taciroglu (USA)Su Tylor (UK)Dan Tobias (USA)Mark Williams (USA)Takashi Yamane (Japan)
Zhiwen Zhu (China)

10Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Plenary Speakers(alphabetical order)

Professor Issam E. Harik, Ph.D.
University of Kentucky, USA
Since joining the University of Kentucky in August 1982, Dr. Harik has directed or codirected
research on more than 120 funded projects. Dr. Harik has directed or is currently directing the
research of 55 Post Docs/visiting professors/scholars, 100 graduate students, 70 undergraduate
students, and 24 high school students. Dr. Harik currently serves as a member of engineeringsocieties and 4 committees; he chaired the Transportation Research Board Committee on
Structural FRP (AFF80) from 2009 to 2015 and is a Council Member of the International Institutefor Fiber Reinforced Polymer Composites in Construction and a member of the International
Advisory Committee - Bridge Engineering Institute. He has organized, co-organized, and chaired
national and international conferences, workshops, and technical sessions. Dr. Harik has
authored or co-authored more than 360 technical publications.

11Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Plenary Speakers(alphabetical order)

Professor John J. Myers, Ph.D., P.E.
Missouri University Science and Technology, USA
Dr. John J. Myers is Professor of Structural Engineering in the department of Civil, Architecturaland Environmental Engineering at Missouri University of Science and Technology and is Deputy
Director of the Missouri Center for Transportation Innovation. He served as Associate Dean for
Academic Affairs in the College of Engineering and Computing from 2015 to 2019. Dr. Myers
has been a member of the Missouri S&T faculty since 1999. He holds Ph.D. and M.S. degrees incivil engineering from the University of Texas at Austin and a bachelor's degree in architectural
engineering from Pennsylvania State University. He previously served as the college's acting
vice provost and dean. Dr. Myers oversees all curriculum and instruction in the College ofEngineering and Computing along with development and implementation of strategic planning
related to academic programs. He establishes plans for faculty development and recruitment aswell as program development and also leads efforts in the development and implementation of
the college's recognition and award structure.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Plenary Speakers(alphabetical order)

Professor Antonio Nanni, Ph.D., P.E.
University of Miami, USA
Dr. Antonio Nanni, President of the American Concrete Institute (ACI), is a structural engineer
interested in construction materials, their structural performance, and field application. Hisinterests are in the field of buildings and civil infrastructure sustainability, monitoring andrenewal. He is the Co-Director of two federally funded centers: NSF Industry/University
Cooperative Research Center (I/UCRC) for the Integration of Composites into Infrastructure(CICI); and, US DOT University Transportation Center on Research on Concrete Applications for
Sustainable Transportation (RE-CAST). In the past 30 years, he has obtained experience inconcrete and advanced composites based systems as the principal investigator of projects
sponsored by federal and state agencies, and private industry. Over the course of this time, hisconstant efforts in materials and structures research have impacted the work of several
technical committees in the US and abroad. Dr. Nanni is the Editor-in-Chief of the ASCE Journalof Materials in Civil Engineering and serves on the editorial board of other technical journals.
He has advised over 60 graduate students pursuing MSc and PhD degrees, and published over
200 and 310 papers in refereed journals and conference proceedings, respectively, in addition
two co-authoring two books. Dr. Nanni has maintained a balance between academic andpractical experience and has received several awards including: 2014 IIFC Medal, International
Institute for FRP in Construction; ASCE 2012 Henry L. Michel Award for Industry Advancementof Research; and, Engineering News-Record Award of Excellence for 1997, (Top 25 Newsmakers
in Construction). He is a registered PE in Italy, FL, PA, MO, and OK.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Plenary Speakers(alphabetical order)

Mr. Takashi Yamane, P.E.
Kyokuto Kowa Corporation, Japan
Takashi Yamane is the President of Kyokuto-Kowa Corporation, a prestressed concrete bridge
construction company in Japan. He has been a registered Professional Engineer in Japan since2006. He graduated from Kure National College of Technology and joined Kyokuto-Kougyo
Corporation (formerly Kyokuto-Kowa Corporation) in 1980. Mr. Yamane earned a master's
degree in civil engineering from the University of Nebraska, Lincoln, in 1995. His work
experience includes design and construction of various types of prestressed concrete bridges,
such as box girder bridges with a balanced cantilever construction, cast-in-place hollow slab
bridges with span-by-span construction, and composite I-section girder bridges. Mr. Yamane led
the engineering department and sales department of the company before becoming president.
He had been a member of the Japan Prestressed Concrete Contractors Association, and chaired
the committee on design guidelines for precast/prestressed concrete splice-girder bridges. He
has published many technical papers and patents in the bridge engineering field.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Plenary Speeches

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

A Review of 30 Years of Experience in Field Testing of Bridges
Issam Harik1*
1: University of Kentucky, Lexington USA; email: Harik@uky.edu*: corresponding author

Keywords: Bridges, Steel, Truss, Concrete, Dynamic, Static, Rating
Abstract: A review is presented of the author and his team’s experience with field testing of
more than 30 bridges in Kentucky. From the late 1980s to late 1990s, long span bridges over the
Ohio and other rivers were primarily tested to evaluate their seismic vulnerability whensubjected to projected earthquakes. Two cable stay bridges were also tested prior to opening totraffic to capture their baseline dynamic properties for comparison with properties derived fromfuture tests. From the mid-1990s to 2022, field tests were primarily conducted on short andmedium span bridges to evaluate the effectiveness of Heavy bridge deck overlays, bridges withand without diaphragms, effectiveness of bridge retrofit, and load rating of bridges. Highlights
of bridge testing and evaluation are presented.
1. Introduction
The 1989 Loma Prieta Earthquake highlighted the seismic risk to double deck highway bridges
with the partial collapse of the San Francisco-Oakland Bay Bridge and the Cypress Viaduct.
Following the earthquake, the Federal Highway Administration (FHWA) commissioned theseismic evaluation of all double-deck bridges located in seismically active regions. Two of thebridges over the Ohio River are double deck bridges, the Brent-Spence Bridge between
Covington, Kentucky (KY) and Cincinnati, Ohio, and the Sherman-Minton Bridge between
Louisville, KY and New Albany, Indiana. Both bridges fall under the category of non-regularbridges and could not be evaluated using the 1996 AASHTO Standard Specifications and the1998 AASHTO-LFRD seismic design and analyses methods. Consequently, more rigorous
assessment is required, and is dependent on the bridge as-built conditions and the creation of
accurate finite element models to assess their behavior during projected seismic events. The
process includes preliminary finite element (FE) modeling, followed by non-destructive dynamic
field testing and calibration of the FE model. The Brent-Spence Bridge (Fig. 1a) is the author’s
first involvement with field testing and evaluation of bridges (Harik et. al., 1997).
In addition to dynamic field testing for seismic evaluation of several long span bridges over the
Ohio, Cumberland, and Tennessee Rivers (e.g., Fig. 1b), dynamic field testing was carried out tocapture the baseline dynamic properties of the Owensboro and Maysville (Fig. 1c) cable staybridges. While, for the Roebling Bridge (Fig. 1d), dynamic field testing was carried out tocalibrate the finite element model used for load rating.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Static load testing was carried out on more than twenty short and medium span bridges toevaluate the effectiveness of the deployment of new design and construction details, e.g.,
removal of intermediate diaphragms (Griffin and Harik, 1997), or load rate the bridge for load
posting (Hutchinson et al., 2021).

(a) Brent-Spence Double Deck Bridge(b) US51 Bridge(c) Maysville Bridge(d) Roebling Bridge

Fig. 1. Bridges over the Ohio River assessed using finite element models calibrated based on
dynamic field testing
2. Dynamic Field Testing for Seismic Evaluation of Bridges
Long span bridges have limited access to the floor beams along the length of the bridge.
Consequently, acceleration measurements are collected by placing accelerometers on the deck.
Since these bridges are main thoroughfares and bridge closure is not an option, one lane of trafficis generally allocated for field testing, and for a brief period of 2 to 3 hours when traffic volume
is low (e.g., Sunday morning). Ambient vibration measurements under traffic and wind induced
excitations are recorded along the bridge at various locations preselected from the preliminary
finite element model of the bridge.

17Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Following testing, field data is analyzed to determine the natural frequencies and mode shapes ofthe bridge. The three-dimensional finite element model of the bridge is calibrated by adjusting
the stiffness of the bridge members until an acceptable match is observed in few natural
frequencies and mode shapes. Then, the calibrated finite element model is used for the seismic
time history response analysis for projected earthquakes. The analysis reveals deficiencies
requiring further evaluation to determine retrofit measures.
2. Static Field Testing for Load Rating
Field testing for load rating has generally been carried out on bridges whose structural plans are
not available. The lack of information about the bridge often yields overly conservative
theoretical load ratings leading to load postings that may exclude school buses and emergency
vehicles from traversing the bridges.
Field testing is carried out using loaded trucks with measured axle weights. Strain gauges, and
LVDTs where applicable, are placed at predetermined stations along the bridge. The data
gathered during the positioning of the truck axles at designated locations along the bridge is used
to load rate the bridge.
3. Conclusions
Field testing has been carried out on more than thirty bridges in Kentucky by the author and his
team since the late 1980s. The primary advantage of testing and evaluation of bridges is the“accurate” determination of the as constructed or current condition of the bridges. Suchknowledge is essential in the evaluation of seismic vulnerabilities, load posting, new materials,
structural systems, etc., and in the decision making on whether to replace or strengthen a bridge,
limit vehicle weights, etc. Such decisions are essential for the safety of the safety of thetransportation system.
4. References

Griffin, J.J. and Harik, I.E., 1997. Experimental Analysis and Analytical Modeling of Bridges with andWithout Diaphragms, 14th International Bridge Conference, Pittsburgh, Pennsylvania, 122-127.

Harik, I.E., D. L. Allen, R. L. Street, Guo, M.W., Graves, R.C., Harrison, J., and Gawry, M. J.
1997. Free and Ambient Vibration of Brent-Spence Bridge, ASCE Journal of StructuralEngineering., 123(9), 1262-1268.
Harik, I.E., D. L. Allen, R. L. Street, Guo, M.W., Graves, R.C., Harrison, J., and Gawry, M. J.
1997. Seismic Evaluation of Brent-Spence Bridge, ASCE Journal of Structural Engineering.,
123(9), 1269-1277.
Hutchinson, C., Peiris, A., and Harik, I.E., 2021. Load Testing and Rating of a Concrete Arch
Bridge, International Conference on Structural Engineering and Construction Management(ICSECM 2021), Kandy, Sri Lanka, 16 pp.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Performance-Based Electrochemical Analysis of an Innovative Magnesium
Phosphate Cement Coating to Mitigate Mild Steel Bar Corrosion
John J. Myers1*, Fan Zhang, and Hongyan Ma
1: Missouri University of Science and Technology, Rolla, Missouri, USA; email: jmyers@mst.edu*: corresponding author

Keywords: magnesium potassium phosphate cement; ribbed steel bars; coating; anti-corrosion;
double protection system.
Abstract: Corrosion of steel is the largest threat to concrete structures including bridges whether
it is a superstructure element or substructure element. In the United States, the total direct cost of
corrosion is estimated at $276 billion per year, which is 3.1% U.S. gross domestic product (GDP).
Preventing and mitigating steel corrosion is critical to ensure long service-life in a RC structure.
This study engages in the investigation of the anti-corrosion performance of rebar in magnesium
phosphate cement (MKPC) coating and ordinary Portland cement (OPC) coating detailed theelectrochemical properties from the open-circuit potential (OCP), polarization resistance (PR) andelectrochemical impedance spectrometer (EIS), and morphology from scanning electron
microscope (SEM). While the MKPC coating holds promise for new construction, the focus of
this paper will be to examine the laboratory results to date and how it may be most suitable forrehabilitation of corrosion damaged mild reinforcing steel often found in older bridge bridges andinfrastructure elements exposed to harsh exterior environments. This study investigated how
MKPC performed compared to traditional OPC and was benchmarked to black steel samples. The
results show that the MPC coating layer offered far better corrosion protection than OPC coatingand hold great promise for a new innovative coating technology to mitigate mild steel corrosion inan economical sustainable fashion.
1. Introduction
Currently, the most efficient and common method to protect mild steel bars in an external corrosive
environment is by building a chemical or physical barrier between the steel and environment.
Improving the anti-corrosion performance of reinforcement in concrete can be achieved bymodifying the ordinary Portland cement’s cementitious protective cover (i.e., using a specific type
cement, adding a super-plasticizer or other admixtures [1] and changing the water-cement ratio, or
by using the cathodic method to protect the bars (i.e., hot-dip galvanized coating [2] andmagnesium alloy anode [3]) or by providing a special coating over bars before casting fresh
concrete, such as fusion bonded epoxy and enamel coating [4]. In addition, another mechanism is
to form a passive film on the steel surfaces to prevent the corrosion process. One of the mostcommon is the passive film formed due to high alkalinity from the ordinary Portland cement
system [5]. However, current methods of preventing corrosion have different limitations or flaws.
Incomplete coating over the entire surface of the bars will lead to small pits by human error, which
affects the long-term performance of corrosion resistance [6-8].

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Magnesium potassium phosphate cement (MKPC) is made of dead-burned magnesium oxide(MgO), the calcination temperature is 1000-1500 ℃ (1832 to 2732 °F), and potassium dihydrogen
phosphate, KH2PO4 (KDP) through an acid-based chemical reaction. Compared to traditional
concrete, MKPC possess many beneficial characteristics, such as high early-strength, favorable
bond strength to old concrete and steel, and a low permeability with good durability characteristics.
Therefore, magnesium potassium phosphate cement holds great promise as a fast repair material
for reinforcement concrete structures due to its fast-setting nature and good steel corrosion
inhibition capacity.
2. Experimental Details
2.1 Materials
A dead burned magnesia (DBM) and a chemical reagent of 99% purity potassium dihydrogen
phosphate (KH2PO4, KDP) were supplied by Martin Marietta Magnesia Special Ties, LLC from
Manistee MI, United States and ICL Specialty Fertilizers-Americas from Summerville SC, UnitedStates, respectively. The Type I Portland cement (PC) supplied by the LafargeHolcim Ltd building
materials company was used to produce specimens coated with ordinary Portland cement paste as
one comparison group. This group was compared with the specimens coated in magnesium
potassium phosphate cement paste.
2.2 Specimen Types, Solution Exposure and Coating Process
In this research, three testing sample conditions were investigated: Group I (UN) is the bare mild
steel bar without any surface treatment. The samples of Group II (OPC) were coated with ordinary
Portland cement paste. Group III (MKPC) is mild steel bar coated with magnesium potassium
phosphate cement paste. The MgO to KDP (M/P) mole ratios and water to binder (w/b) ratios of
the MKPC paste using in the research are 6 and 0.18, respectively, and the retarder was not usedin this research. Since the compressive strength of MKPC paste was about 27.58 MPa (4000 psi)
at the seven days testing age, the same compressive strength for OPC paste was needed, thus, the0.38 w/b ratio for OPC paste was used in this research.
The 3.5 wt.% NaCl corrosion solution was made by mixing purified sodium chloride with
deionized water, so it means that the chloride concentration is about 0.6 mol/L. After magnetic
stirring in an indoor environment [temperature is 20±2 ℃ (64 to 72 °F) and air humidity is
55±10%], all samples were immersed into this solution. Before the electrochemical test,
conductivity and temperature of the 3.5 wt.% NaCl corrosion solution were determined by a Hanna
Instruments HI5522 Meter. These two values stabilize at 55.0±5 mS/cm and 20±2 ℃ (64 to 72 °F),
respectively. During the entire EC test period, the conductivity and temperature of the corrosion
solution was measured in the same way, the values kept stable, and no changes occurred.
Furthermore, during the EC test, the pH value of corrosion solution also was monitored.
3. Test Methods
The electrochemical measurements and data were performed and collected by using a typical threeelectrode setup. A 25.4 x 25.4 x 0.254 mm (1 x 1 x 0.01 in.) platinum sheet, a saturated calomelelectrode (SCE) and the prepared specimens, as shown in Fig. 1, were used as counter electrode(CE), reference electrode (RE) and working electrode (WE), respectively. The EC test setup is
shown in Fig. 2.

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Fig. 1. Geometry of rebar samples (unit: mm)

Fig. 2. Three-electrode setup

In addition to the EC testing, specimens underwent the SEM test using a Helios Nanolab 600
coupled to an energy dispersive X-ray spectroscopy (EDS), which was used to analyze the
distribution of chemical elements of the cross section of specimens from the inner substrate mild
steel part to outside coating part. The surfaces of substrate steel of MKPC specimens underwent
the XPS test to analyze the chemical group formed due to the reaction between MKPC paste andmild steel rebar. The TG/DTA test was conducted on the prepared treated powder by using a TA
instruments SDT Q600 under nitrogen atmosphere. To analyze the chemical composition of
cementitious coating, X-ray diffraction (XRD) data of prepared specimens were collected by a
Panalytical X’pert Pro MPD diffractometer.
4. Results and Discussion
Due to length requirements of the paper, highlights of the above-mentioned experimental tests will
only be presented. Fig. 3 shows the corrosion maps of three sample types, which were immersed
into 3.5 wt.% NaCl corrosion solution and tested under the same indoor environment, after the5376 hours (224 days) electrochemical test. In these figures, the degree of corrosion trends tobecome weak from the samples without coating compared to the samples with ordinary Portland
cement coating, and lastly, to the samples with magnesium potassium phosphate cement paste
coating.

Fig. 3. Bar surface conditions of samples after 5376 hours EC test for:(a) UN, (b) OPC and (c)MKPC.
Table 1 shows the values of corrosion current density (Icorr) and corrosion potential (Ecorr) whichwere extracted from potentiodynamic polarization plots, and the corrosion rate (CR) was
calculated based on (1):21

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023
𝐼𝐼 ∙𝐾𝐾∙𝐸𝐸𝐸𝐸

𝐶𝐶𝐶𝐶 = 𝑐𝑐

(1)

𝑑𝑑∙𝐴𝐴

where CR is corrosion rate, Ic is the corrosion current in Ampere, K is the constant that defines theunits of the corrosion rate, EW is equivalent weight (in g/equivalent), d is the density, and A is the
sample area. This shows the significant magnitude of order in corrosion rate reductions.
Table 1. Parameters from potentiodynamic polarization curves for OPC and MKPC specimens.
Ecorr (mV/SCE)

Icorr (µA/cm2)

Corrosion rate (mm/year)

Time(hours)

OPC

MKPC

OPC

MKPC

OPC

MKPC

5

-868.568-997.484

14.7637.903

0.17128

0.091691344-953.089-969.436

27.4535.971

0.318510.069285376-825.883-999.615

27.051

2.051

0.313850.02379

5. Concluding Remarks
This study was developed to investigate and evaluate the steel bar anti-corrosion performance
under the protection of the ordinary Portland cement paste and magnesium phosphate cement paste
coating referred to as OPC and MKPC, respectively in this work.
• The MKPC coating impedance kept increasing until the 3369 hours (140 days) andmaintained stability to the end of test in the corrosive environment-increasing around 7
times from the first test, supporting that the MKPC coating paste developed a far better
excellent anti-corrosion performance than OPC coating.
• The MKPC samples’ integral coating demonstrated no flaws after 5376 hours (224 days),
therefore the magnesium potassium phosphate cement paste coating displayed strongrobust properties to delay the corrosion rate.
• The MKPC material may be considered to be practical for repair as well some underwater
construction and some traditional reinforcement concrete structures to reduce the corrosion
rate and to eventually extend service life of the structure.
6. References
1Shi, X., et al., Strength and corrosion properties of Portland cement mortar and concrete withmineral admixtures. Construction and Building Materials, 2011. 25(8): p. 3245-3256.
2

Sistonen, E., A. Cwirzen, and J. Puttonen, Corrosion mechanism of hot-dip galvanised
reinforcement bar in cracked concrete. Corrosion Science, 2008. 50(12): p. 3416-3428.

3

Parthiban, G., et al., Cathodic protection of steel in concrete using magnesium alloy anode.
Corrosion Science, 2008. 50(12): p. 3329-3335.

4Tang, F., G. Chen, and R.K. Brow, Chloride-induced corrosion mechanism and rate of enameland epoxy-coated deformed steel bars embedded in mortar. Cement and Concrete Research,2016. 82: p. 58-73.

5Feng, X., et al., The degradation of passive film on carbon steel in concrete pore solution undercompressive and tensile stresses. Electrochimica Acta, 2011. 58: p. 258-263.

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6

Behzadnasab, M., S. Mirabedini, and M. Esfandeh, Corrosion protection of steel by epoxy
nanocomposite coatings containing various combinations of clay and nanoparticulate zirconia.
Corrosion Science, 2013. 75: p. 134-141.
7
Tang, F., et al., Corrosion resistance and mechanism of steel rebar coated with three types ofenamel. Corrosion Science, 2012. 59: p. 157-168.
8

Etteyeb, N., et al., Corrosion protection of steel reinforcement by a pretreatment in phosphate
solutions: assessment of passivity by electrochemical techniques. Corrosion engineering, scienceand technology, 2006. 41(4): p. 336-341.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Diagnostic Load Testing of a Bridge Before and After Rehabilitation
Jiayi Ding1, Nicolas Begasse2, Adriana Toro3, Francisco De Caso4, and Antonio Nanni5*
1: HDR, Doral, FL, USA; email: jiayi.ding@hdrinc.com: HDR, Doral, FL, USA; email: nicolas.begassededhaem@hdrinc.com
3: Broward County, FL, USA; email: atoro@broward.org
4: University of Miami, Coral Gables, FL, USA; email: fdecaso@miami.edu
5: University of Miami, Coral Gables, FL, USA; email: nanni@miami.edu
2*: corresponding author

Keywords: Sonovoid bridge; deterioration; load testing; rehabilitation technique; serviceability
Abstract: This paper provides the experimental results of two diagnostic load tests conducted ona Sonovoid® (i.e., prestressed concrete voided slab) bridge in Broward County, FL, which wasbuilt in the late 80’s. Diagnostic load tests were conducted pre- and post-rehabilitation toevaluate the bridge's behavior and effectiveness of rehabilitation technique. The collected data
was utilized to determine the lateral load distribution, both before and after rehabilitation. Actual
live load distribution factors were calculated based on deflection and strain measurements.
Among the data collected were: a) concrete longitudinal strains and vertical deformations of slab
units within the tested span; and, b) strains in dowel bars forming part of the repairs. Diagnostic
load testing also provides the decision-making tool for evaluating this repair technique forsimilar bridges in the future. Considering that most of this rehabilitation work was performed
under the bridge, the need for lane closure is significantly reduced. Due to the novelty of therehabilitation technique, load testing was necessary to evaluate its effectiveness.
1. Introduction
Diagnostic load tests (Olaszek et al., 2014) are the most common type of testing techniques usedin new bridges as well as existing ones. The load test method can be used to check several in-situ
parameters for a given bridge (Lantsoght et al., 2017) as well as the verification of the design
assumptions and analytical models used for design (AASHTO 2017). The verification of thedesign assumptions can be carried out by comparing the measured structural response and theanalytically determined response. Two diagnostic load tests were conducted on a Sonovoid®
bridge to evaluate the behavior characteristics of the bridge under static and dynamic (moving)loads, both before and after rehabilitation. The load tests aimed to determine the load distribution
between the Sonovoid® beam elements spanning a segment of the bridge and provide
experimental data for modeling and verification purposes.
Sonovoid® bridge. The bridge, located in Broward County, FL, is a five-span structure, withnorth and south bound traffic and pedestrian sidewalks, where each span is 36-ft. (10.97-m) longas seen in Fig. 1(a). The spanning elements are prestressed concrete hollow Sonovoid® slab
elements as seen in Fig. 1(b). The north and south bound superstructures are separate structuresconnected only by a raised median slab. At all other spans, the superstructure is one continuous
structure. The test Span 4 (south bound) consists of a total of 12 Sonovoid® elements.

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The rehabilitation technique utilizes transverse Near-Surface-Mounted (NSM) stainless steel
dowels with Carbon-Fiber-Reinforced-Polymers (CFRP) laminates. Each dowel bar is inserted in
a 1¼ x 1¼ in. (32 x 32 mm) groove and filled with epoxy. CFRP laminates are then provided on
the bottom face of slab elements over the dowels. This technique was selected for further
investigation because of its potential efficiency, lower cost, and minor impact to traffic during
construction. The enhanced live load distribution would result in less flexural demand on theindividual slab units. In addition, enhancing the interconnection between adjacent units would
mitigate relative movement, and ultimately prevent the existing shear keys from further cracking
and deterioration. Externally bonded CFRP laminates were applied to support the concrete coveragainst the peeling forces induced by the dowel steel bars. In addition, CFRP is intended toprovide an enhanced level of protection against possible corrosion of the steel bars.

(a)

(b)
Fig. 1. Sonovoid® bridge, located in Broward County, FL: (a) aerial view, and (b) typical unit
section (1 in.=25.4 mm; 1 ft. = 305 mm)
Instrumentation. Instruments used in the load tests were: four spring-loaded Linear Variable
Differential Transformers (LVDTs); 12 spring-return linear potentiometers: eight spring-loaded
dial gauges with an internal closed-circuit television (CCTV) used to visually monitor
deflections real time; two inclinometers with integrated sensor electronics; 14 wire strain gauges
bonded to concrete; 10 rosette gauges bonded to steel dowels. A dedicated suspended access
scaffolding with a rigid support frame system was installed under the bridge to hold and secure
all instruments during the load tests (See Fig. 2). The scaffolding system was fastened to the bent
caps located on either side of Span 4, providing a means of accessing the underside of the bridge
for instrument installation, maintenance, and removal before, during, and after testing. To enable
this, a floating platform over the water was utilized.

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Load Protocol. Six Load Cases (LCs) were considered as part of load protocol for the bridge
Load Test 1 (pre-rehabilitation), while seven LCs were considered for Load Test 2 (postrehabilitation). For each LC, the same fully-loaded truck with gross weight of 67,440 lb. (300kN) was used. For the static load tests of each LC, the truck remained stationary on Span 4 withdouble rear axles placed on either side of the mid-span line. For the dynamic load tests, the truck
moved over the bridge in the south bound direction at an approximate constant speed of 10 mph(16 kph), driving over a 4 x 4 in. (100 x 100 mm) piece of lumber placed perpendicular to thetravel direction. Two load repetitions per each LC and load protocol were performed.

Fig. 2. Overall view of instruments on the scaffolding
2. Results
The results herein reported are based on average net values (absolute magnitude) for the two
repetitions. The average net measurements are computed by the difference between the initial
reading and the average maximum value recorded at the time of load application.
Linear Variable Displacement Transducers (LVDTs). Only LVDT readings for Load Test 1(before repairs) are summarized in and Table 1 for the dynamic load protocol. LVDTs measured
the vertical (LV2 and LV4) and horizontal (LV1 and LV3) mid-span displacements (deflections)of slab units No. 4 and No. 10, respectively (numbered from the west side).
Table 1. LVDT mid-span displacements for the dynamic load for Load Test 1(1 in.= 25.4 mm)
Equipment ID
LV2 (Vertical)
LV4 (Vertical)
LV1 (Horizontal)
LV3 (Horizontal)

Unit
in.
in.

Slab #
4
10
4
10

LC1D
0.144
0.025
0.012
0.014

Dynamic Load Protocol
LC2D
LC3D
LC4D
LC5D
0.193
0.147
0.100
0.084
0.050
0.087
0.134
0.182
0.011
0.011
0.011
0.011
0.025
0.028
0.021
0.020

LC6D
0.053
0.219
0.012
0.018

Potentiometer: Static Load Tests. Fig. 3 shows the vertical displacement differences between
Load Test 1 and Load Test 2 under LC2 providing a representation of the relative shear transfer
between the slab units.
Potentiometer: Dynamic Load Tests. Deformation results of dynamic load protocol were basedon maximum readings of each Load Case. Results like those shown in Fig. 3 were obtained forvertical displacement differences between Load Test 1 and Load Test 2.

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Fig. 3. Difference between Load Tests 1 and 2 for static LC2 (1 in. = 25.4 mm)
3. Discussion• The results of the first and second repetitions across various LCs and load protocols
demonstrated a high degree of repeatability with minimal differences.
• Dial gauges and CCTV effectively monitored the bridge responses and validated verticaldisplacements during load tests.
• Horizontal displacements were relatively constant and negligible, across the different Load
Cases and load protocols. Vertical displacements correlated directly to the position of thetruck and Load Case, where there was a relatively high transfer of shear load, based on net
deflections (about 75% of the peak displacement) to the adjacent unloaded slab units.
• Dynamic loading imposed higher demand on slab units with varying levels of transfer
depending on the location of the applied load.
• The dynamic load protocol resulted in higher peak displacements, on average 50% morethan under static loads. Similarly, a relatively higher transfer of shear loads, based on net
deflections to adjacent unloaded slab units were observed.
• Longitudinal concrete strain responses were effectively recorded. Strains measured underdynamic load protocol were on average 60% more than under static loads.
• Slab unit No. 4 rotations correlate closely to measured mid-span deformations. This
indicates that the supports of Slab unit No. 4 have limited restraints.
• Composite action between the stainless-steel dowels and Sonovoid® elements was
confirmed, with reduced total magnitude of deflections compared to pre-rehabilitation.
4. Conclusions
This paper provides the experimental results of the diagnostic bridge pre- and post- rehabilitationload tests conducted on a segment of the Sonovoid® bridge. Tests consisted of combining six toseven different Load Cases by positioning a fully loaded truck at different locations on bridge
span No. 4 (South Bound) engaging different slab elements, with two repetitions for each loadcase, and two different load protocols (static and dynamic). Overall, the two diagnostic load tests
proved to be a highly effective tool for assessing the degree of success of the chosen
rehabilitation intervention. By providing the owner and design team with quantitative data and abenchmark for future evaluations.
5. Acknowledgements
The authors would like to thank NSF for partial financial support under grant No. 1916342.

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6. References
AASHTO. 2017. “Manual for Condition Evaluation of Bridges,” AASHTO, 3rd Edition
Washington, DC
Lantsoght, E., der Veen, C., de Boerd, A., Hordijk, D.A. 2017. Bridge engineering instituteconference, Journal of Bridge Engineering Institute, 1(2), 36-45.

Olaszek, Z., Lagoda, M., and Casas, J.R. 2014. “Diagnostic Load Testing and Assessment of
Existing Bridges: Example of Application,” Structure and Infrastructure Engineering, 10, 834-842

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Current State of Expressways and
New Technologies for Bridge Deck Maintenance in Japan
Takashi Yamane1*

1: Kyokuto Kowa Corporation, Hiroshima, Japan; email: tyamane@kkn.co.jp*: corresponding author

Keywords: Precast concrete slabs; Joint; Loop joint; ELSS joint; Productivity; Slab replacement
Abstract: This paper outlines the service and damage status of expressways in Japan, as well as
repair and renewal technologies. The decks of expressway bridges in Japan have been found to be
severely damaged, and major renewal projects centered on deck repair and renewal are therefore
underway. One challenge in these projects is shortening the traffic restriction period. Therefore,
this paper outlines three technologies that can shorten this period when repairing/renewing a bridgedeck. The first is a technology for removing an existing deck; the second is a technology forbonding PC slabs when renewing a deck; and the third is a preventive maintenance technologythat does not require any traffic restrictions.
1. Introduction
In Japan, where the popularization of the automobile occurred later than in the US and Europe,
expressway construction also began later. The first expressway opened in Japan in 1963.
Construction has progressed rapidly since then, and an expressway network extending overapproximately 9000 km is in place today. Approximately 3700 km (roughly 40%) of this network
consists of roads that have been in service for 30 years or more (Fig.1).

Fig. 1. Expressway network in Japan

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When we look at the expressways by type of bridge structure, prestressed concrete bridges accountfor 35%, reinforced concrete bridges for 32%, and steel bridges for 33%.
Shifting our focus to the damage status of expressway bridges, the most severe damage isto the reinforced concrete decks of steel bridges. Survey results show that the prominent causes
are salt damage and heavy traffic (Fig. 2).

Fig. 2. Cause of damage of expressway bridges
2. Outline of Renewal ProjectIn 2014, the three companies that manage Japan’s main expressway network (East Nippon
Expressway Company Limited, Central Nippon Expressway Company Limited, and West Nippon
Expressway Company Limited) designed major renewal and repair projects and embarked on theirimplementation. The scale of these projects is extremely large, with the combined cost of the major
renewal and repair projects totaling $22,271,000,000. Of this amount, deck slab replacement work
accounts for approximately 55%, or $12,170,000,000.
3. New Technologies for Bridge Deck Maintenance
3.1 Deck removal for composite girder bridge
Steel bridges include composite and non-composite structures. In composite structures, numerous
studs are welded to the top flanges of the steel girders, making deck removal difficult and labor
intensive. The technology introduced here is for removing the concrete deck slab of a composite
steel girder bridge by cutting it into large blocks.
First, a cutter is used to divide the existing slab into blocks that can be lifted with a crane.
Next, an electric wall saw is used to cut the middle of the studs from the top surface of the slab
(Fig. 3).

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Fig. 3. Electric wall saw to cut the studs
This separates the stud heads from the steel girder, so there is very little resistance when the slab
is pulled off the girder. Then, the slab is jacked up and removed in block units (Fig. 4). This
technology can drastically reduce on-site labor and shorten the work period.

Fig. 4. Jack up block units
3.2 Panel to panel connection
This technology is applicable to works in which a bridge deck is constructed by connecting
precast/prestressed concrete slabs on site. It enables connections between precast slabs, whichconventionally have a reinforced concrete construction, to be made using only a filling material
that is softer than the slab concrete (We call this system “ELLS Joint.”). Safety, durability,
drivability, and economic efficiency were confirmed by analyses and full-scale experiments. The
deck is supported continuously in the driving direction by the steel girders and the wall balustrades.
When this kind of deck is constructed using precast slabs, a filling material that is less rigid thanthe main body is used to fill the joints, so that the deck becomes an anisotropic slab. As a result,the bending moment in the driving direction is reduced, while the bending moment in the roadwidth direction is increased. In response to this, the required strength is secured by increasing thenumber of PC steel bars in the precast slab (Fig. 5). Employing this joint structure results in
reductions of 10% or more in manpower, number of workdays, and cost.

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Standard practice
Newly developed joint
Fig. 5. Panel-to-panel connections
3.3 Lithium pressurized injection for preventive maintenance
Lithium pressurized injection is a technology that involves injecting a lithium nitrite solution athigh pressure into the concrete. This technology was developed to inhibit internal expansion of
concrete associated with the alkali-silica reaction by taking advantage of the nature of lithium ions.
However, because the nitrite ions reach the rebar, this technology also has the effect of preventing
rust on the reinforcing steel. Making use of the latter effect, rebar corrosion in the upper and lower
surfaces of a deck slab can be prevented by injecting a lithium nitrite solution from the undersideof the deck. Because all the work is done from the underside, this method does not involve traffic
restrictions (Fig. 6). This is a very useful advantage in bridge maintenance, and we hope to expand
the applications of this method in the future (Fig. 7).

Fig. 6. Lithium pressurized injection from underneath the deck

Fig. 7. Maintenance work without traffic regulation

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4. Conclusion
All three of the new technologies introduced in this paper have a proven track record in Japan.
However, there are issues such as demonstrating effectiveness and improving work efficiency. We
intend to improve the level of performance of these technologies by continuing to conduct
experimental studies and on-site demonstrations.
5. Acknowledgments
We would like to take this opportunity to thank Dr. Yasushi Kamihigashi of the Central Nippon
Expressway Company Limited who provided us with information about Japan's expressways.

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Materials and Performance

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Quality of Lamination Structure Made by 3D Printing under Different
Lamination Conditions
Yuji Yamada1*, Tomoko Fukuyama2, Yunmi Kim3,
Hiroharu Kamada4, and Dhruva N. Katpady 5
1: Yamaguchi University, Yamaguchi, Japan; email: y-yamada@yamaguchi-u.ac.jp: Ritsumeikan University, Shiga, Japan; email: tmkfkym@fc.ritsumei.ac.jp
3: Hanbat National University, Daejeon, South Korea; email: kymmiss123@naver.com
4: Polyuse, Kanagawa, Japan; email: hiroharu.kamada@polyuse.xyz
5: Fukuoka University, Fukuoka, Japan; email: dhruva@fukuoka-u.ac.jp
2

Keywords: 3D printing technology; Quality of interlayer; Lamination conditions
Abstract: In this study, specimens of lamination structure were made by 3D printing using rapid
hardening mortar. In addition, lamination time and season were changed when made thespecimens. In order to evaluate the quality of joining part, these specimens were investigatedabout the mechanical properties, the shrinkage properties and the air permeability.
1. Introduction
Recently, the 3D printing technology is attracting attention as a construction method which canachieve an automated construction. In the 3D printing technology, mortar is used to makefilaments, and structures are made by laminating the filaments. However, there is concern about
durability of interlayer in the lamination structure. Also, in Japan, there are extreme climaticchanges throughout the year. Therefore, in the actual construction, the environmental effects
such as temperature and moisture may affect the quality of interlayer in a lamination structure.
In addition, mortar which is used to lamination by 3D printing has a characteristic of rapid
hardening. Therefore, an usable time of mortar for good lamination is short compared withnormal concrete, and it makes difficult to obtain the high durability. In this study, the specimensof lamination structure using rapid hardening mortar were made. In addition, lamination time andseason were changed when made specimens. Also, these specimens were evaluated aboutmechanical properties, shrinkage properties and mass transfer resistance.
2. Experimental Outline
2.1 Mix proportion of mortar
Prepacked mortar which is generally used as a material of patchwork repair was used in purposeof making filaments. Conditions of mix proportion were W/C=36%, S/C=1.3. In addition,polyamide fiber (length: 10mm) was mixed at 0.2 % (vs. total volume of mortal), and also
expansive materials (type of CSA) was included. In order to get the rapid hardening, polymer
was mixed.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Frame

Hose
Nozzle(a)(b)
Photo. 1 Outline of 3DP specimen: (a) 3D printing equipment; (b) Overview of specimens
Strain gage

Dataloggar150mm

Position of
air permeability test

Polished elaborately

CylinderPiston is fixed．

Putty

Negative pressure

Specimen

150mm

Low er layer Upper layer

3DP smecimen

After 2 months

Fine pressure
difference meterAir flow s in cylinder
300mm

(b)

(a)

Fig. 1 Outlines of experiment: (a) Drying shrinkage test; (b) Air permeability test

2.2 Outline of Specimens
Outline of specimen is shown in Fig. 1. The specimens (3DP specimens) were made by 3D
printing, as shown in Photo. 1. The lamination time between upper layer and lower layer waschanged (0min. or 30min.). The specimens were made in summer (June) and autumn (October).
After that, these specimens were exposed outside (environmental condition was general atmosphere) for 2 months (summer season specimen: June-August, autumn season specimen:
October-December).
2.3 Test Methods(1) Setting test of mortar
Setting test of mortar was conducted conforms to Japanese Industrial Standard (JIS A 1147).
(2) Mechanical Properties test
Compressive strength test was conducted. The specimens were cylinder (φ: 50mm, height:
100mm) . The curing conditions were same with the above exposure conditions.
(3) Drying shrinkage test
The exposure specimens had been polished surface, and then those specimens were attachedstrain gages and measured the strain, as shown in Fig. 1 (a).

36

0.3
SummerAutumn

0.25
0.2
0.15
0.1
0.05
0

0

10

20

30
40
Time (min.)50

60

Fig. 2 Setting test results

500

Upper layer
Lower layer

400
300

Summer

200
100
0
-100
-200

0

7

14

21 28 35 42
Period (day)

49

56

(a)

500
400
300
Strain

3. Experimental Result and Discussion
3.1 Setting test results
The setting test results are shown in Fig. 2. Where,
the air temperatures were the case of summer
season: 30.1 ℃ and the case of Autumn season:
23.5 ℃ , mortar temperature were the case ofsummer season: 28.9℃ and the case of autumn
season: 22.3℃. The penetration resistance valueof summer season was larger than that of autumn
season. However, regardless of season, the time
when penetration resistance value achieves
0.1N/mm2 were less than 30 minutes. According
to the research of the past (Okazawa andSugamata, 2001; Matsushita and Sue, 2001), in
order to prevent the cold-joint in construction ofa concrete structure, the placing joint timeinterval must be shorter than the time which
achieves the penetration resistance value =0.1N/mm2. It might be also similar in case of
constructions by 3D printing technology too.

Strain(4) Air permeability test
After the drying shrinkage test, exposure
specimens were conducted the air permeability
test by cylinder method (Watanabe et al. 2018).
Outline of test method is shown in Fig. 1 (b). The
joining part between a specimen and a syringe
were sealed by a putty. The piston was pulled up
and then fixed during 300second. variationpressure internal syringe was measured by a finepressure difference meter.

Penetration resistance value (N/mm 2)

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

200

Upper layer

AutumnLower layer

3.2 Results of drying shrinkage test
100
An example of the results in drying shrinkage
test are shown in Fig. 3. The lamination time0interval is 30minutes. The margins of the strain-100
between the upper layer and the lower layer were-200
large in case of autumn. The compressive0
7 14 21 28 35 42 49 56
strength of mortar were autumn season:
Period (day)22
40.2N/mm and summer season: 55.2N/mm .
(b)
Therefore, the bonding strength between theupper layer and the lower layer of specimen
Fig. 3 Results of drying shrinkage test:
which had been made in autumn season might be(a) results of summer; (b) results of autumn
lower than that of summer. In addition, theautumn environment might be occurred thedrying shrinkage easily compared with the summer environment due to the low moisture. From

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

the above, these results assume to effected by multiple cause including the physical properties ofmaterial and the environmental conditions.
Time (sec.)

Inner pressure of syringe (kPa)

3.3 Results of air permeability test
The results of the air permeability test are shownin Fig. 4. Where, the legends which are shown in
Fig. 4 are “Season (Summer: S, Autumn: A)” “lamination time interval (min.)”. A-0 and S-0
had the high airtightness, because the pressure
fluctuations had been low. On the other hands,
the pressure fluctuations of A-30 and S-30 had
been striking. Therefore, A-30 and S-30 had the
low airtightness compared with A-0 or S-0. In
particular, the airtightness of A-30 was mostinferior, because the pressure of A-30 rose fasterthan that of S-30. These results have acorrelation with the results of the drying
shrinkage test.

0

0

50

100

150

200

250

300

-15
-30
-45
-60

A-0
A-30
S-0
S-30

-75
-90

Fig. 4 Results of air permeability test

4. Summary
From the results of the drying shrinkage test and the air permeability test, the quality of thelamination gap was degraded, due to the time interval of lamination became the time which
achieved at the penetration resistance value = 0.1N/mm2. Also, a degree of deterioration was
different, it was effected by multiple cause including the physical properties of material and theenvironmental conditions.
5. References
Satoshi Okazawa and Takumi Sugamata. 2001. Evaluation for part of placing on consolidated
fresh concrete, Japan Concrete Institute, Concrete Journal, 39 (5), 83-88.
Hiromichi Matsushita and Yoshihiro Sue. 2001. A Study on Relationship between Condition ofPlacing on Consolidated Fresh Concrete and Occurrence of Cold Joint, Journal of Society ofMaterials Science, Japan, 50(8), 851-856.
Takeshi Watanabe, Kenjiroh Omoya, Kenta Kotani, Chikanori Hashimoto and Masayuki
Sekigawa. 2018. Evaluation of Air Permeability of Cover Concrete by Simple Test Method
Using Syringe, The 6th Japan-US NDT Symposium Emerging NDE Capabilities for a Safer
World (Japan-US 2018).

38

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Bond Strength Evaluation of Grout Materials for Closure Pour of Precast
Concrete Bridge Deck Panels
Zhifu Yang1
1: School of Concrete and Construction Management, Middle Tennessee State University,
Murfreesboro, Tennessee, USA; Zhifu.Yang@mtsu.edu
Keywords: Bond Strength; Grout Materials; Substrate Moisture; Curing temperature
Abstract: The use of full-depth precast concrete panels for rehabilitation of bridge decks allows
fast installation and all-weather construction. The panel joints that are connected later with grout
materials may de-bond under load, leading to a less desirable composite action. As a result,testing the bond capacity of various grout materials is essential to establish the quality and theserviceability of the closure pour. In this study, various prepackaged grout products(cementitious and epoxy) across US were requested and their bond strength was evaluated using
flexural and slant shear bond tests. Different substrate moisture conditions (wet, SSD, and dry)as well as different curing temperature (23°C, 30°C, and 35°C) were used to prepare thecomposite specimens. Additionally, steel fibers were added to the grout materials and theireffects on the bond capacity were investigated. It was found that most grout products showed
acceptable to good bond capacity. Specifically, some products exhibited excellent bond strength.
The surface moisture of substrate noticeably affected the bond capacity of grout products.
However, different products showed different effects. Similarly, the curing temperature hadgreat impacts on the bond strength. An increase in curing temperature mostly reduced the bondstrength. Only a few products showed an increased bond strength at a higher curing temperature.
Interestingly, for grout products with high bond strength, adding 3% steel fibers normally caused
a reduced bond strength, while it significantly improved the bond strength for grout productswith low bond strength.
1. Introduction
Conventionally, cementitious grouts, epoxy mortars or polymer concretes are used for the
closure pour. While cementitious grouts are easier to mix and more compatible with thesubstrate; they are more prone to shrinkage. Oppositely, epoxy mortars or polymer concretes
have extremely low permeability and dry shrinkage, as well as good adhesion with the substrate;
but they are very sensitive to moisture variations and thermally incompatible with the substrate.
As a result, testing of various types of grout materials under different service conditions isessential to establish the quality and serviceability of materials.
2. Materials, Proportions, and Experimental Programs
Twenty-five grout materials were collected from 16 different manufacturers. All products wereprepackaged in 23-27kg bags or buckets.
Twenty-three products were cementitious
mortars/concretes and two products (#8 and #18) were epoxy-based mortars. The materials weremixed in a 0.06 m3 rotating drum mixer or in a bucket using a paddle mixer with a typical batch
size of 25-32kg. Water was first added to the mixer or bucket and then the grout material was

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

introduced. An average water content recommended by the manufacturer was used, which
typically resulted in a flowable mixture. The mixture proportion for the substrate concrete was:type I cement: water: 9.5mm crushed limestone: natural river sand = 1: 0.4: 2.43: 1.35. A waterreducing admixture and an air-entraining admixture were used to achieve a slump of 102 to
127mm, and a fresh air content of approximately 6.5%. The 28-day compressive strength of thisconcrete using 102x203mm cylindrical specimens was approximately 38.6 MPa. The substrateconcrete was cured for 28 days, the fresh grout was poured into the mold to achieve thecomposite specimen. The flexural bond test was conducted on the composite specimens
following ASTM C78 and the test setup is shown in Figure 1. The slant shear bond test was
performed following the procedure similar to the one described in ASTM C882. The test setup
is shown in Figure 1.
Flexural Loads
Conventional
Concrete

Grout Material
6"

6"

6"

6"

Fig. 1. Bond test setup

3. Results and Analysis
3.1 Effects of Substrate Surface Moisture on Bond Strength
The effects of substrate surface moisture on the bond strength and the failure plane aresummarized in Table 1. The flexural bond strength varied widely with the substrate surface
moisture conditions. Some grouts exhibited high flexural bond strength under all substratesurface moisture conditions; while some grouts including epoxy grouts, preferred dry substratesurface and some grouts favored wet or SSD substrate surface. When the moisture changed from
SSD to wet, six products displayed a noticeable decrease in the flexural bond strength, thirteen
products showed essentially insignificant changes in the flexural bond strength, and five products
demonstrated a substantial increase in the flexural bond strength. Also, the failure plane did notchange for most products.
These results indicated that a wet substrate surface did notnecessarily reduce the flexural bond strength. When the substrate surface moisture changed from
SSD to dry, twelve products exhibited a reduced flexural bond strength, six products showedroughly equal flexural bond strength, and seven products even demonstrated a significant
increase in the flexural bond strength. In addition, a slight increase in the interfacial failure wasobserved. A comparison of all these results suggested that a dry substrate would be more likely
to reduce the flexural bond strength and cause a higher risk of debonding than a wet substrate.
The effect of substrate surface moisture on the flexural bond strength also depended on thenature of the product. For examples, some products normally failed at the substrate at allmoisture conditions. These grouts typically had high compressive strength and flowability. For
low-strength grouts, failure normally occurred at the grout regardless of the moisture condition.
For some grouts, the specimens normally failed along the interface despite of the moisture
condition, indicating that these materials may be less penetrative. For the two epoxy grouts, dry

40

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

substrate typically showed better bonding and failure normally occurred at the substrate, whereas,
wet or SSD substrates generally led to reduced bond capacity and resulted in the interfacial
failure.
For the slant shear bond test, some products displayed almost equal slant shear bond strength at
all three moisture conditions. Some products showed the highest slant shear bond strength with
SSD, wet, or dry moisture condition. In general, ten products showed a noticeable reduction in
the slant shear bond strength as the substrate surface moisture changed from SSD to wet. Six
products demonstrated approximately the equal slant shear bond strength and eight products evendisplayed a significant increase in the slant shear bond strength. This again indicated that a wet
substrate did not always reduce the slant shear bond strength. Similarly, when the substratesurface moisture changed from SSD to dry, ten products demonstrated a substantial increase,four products showed insignificant variation in the slant shear bond strength, and only ten
products exhibited a reduced slant shear bond strength. This also indicated that a dry substratesurface did not always reduce the slant shear bond strength.
Table 1. Bond strength and failure plane (FP) of specimens with different substrate moisture.
Product
ID
#1
#2
#3
#4
#5
#6
#7
#8
#9
#10
#11
#12
#13
#14
#15
#16
#17
#18
#19
#20
#21
#22
#23
#24
#25

Dry
MPa
FP

2.8
5.6
4.0
2.6
3.3
4.9
0.2
3.7
1.2
4.1
3.1
1.3
1.7
2.4
4.6
4.4
3.1
3.5
0.8
0.0*
3.8
1.3
0.6
0.6
0.0*

I
S
GSI
I
GS
G
I
S
I
G
GS
I
I
GI
G
IS
GI
S
I
I
I
I
I
I
I

MPa

SSD

FP

Wet
MPa
FP

4.1

IS

4.1

I

5.7

S

4.8

S

5.2

IGS

**

**

16.5

I

19.6
16.7

S

4.4
5.2

S

11.5
21.1

G

4.9

G

19.2

GI
I
G

1.0
2.8
3.8

GI
I
G

15.1
20.9
18.6

4.1

G

4.1

G

16.7

2.7
2.1

S
S
I
I
G
SI
IG
IS
GS
I
I
I
I
I
I

2.8
2.1

G
G
G
**
G
GI
S
I
G
I
I
I
I

3.9
1.8
2.6
3.6

0.6
2.3
2.9
4.9
4.4
2.5
1.8
2.2
2.9
1.5
1.0
0.8
0.0*

0.9
1.6
3.8
3.6
4.1
2.2
1.8
0.0*
2.0
2.8
1.8
0.5
0.0*

I
I

SSD

FP

G

4.5
4.6

I

Dry

MPa

18.0
19.5
9.3
18.8
14.4
18.9
19.0
17.4
6.8
*
20.8
22.6
21.5
9.4
2.8

S
IGS
I
GI
G
GIS
S
SI
IS
GIS
IS
I
IG
SIG
IS
S
S
I
I
ISG
S
I
I
I

Wet

MPa

FP

MPa

FP

19.1

S

20.7

S

22.3

S

23.3

S

20.2

IGS

20.1

ISG
SI

I

I

10.4
26.0

IGS

17.8
17.9

22.2

GS

18.4

G

10.0
6.9
18.6

GI
IG
S

7.0
17.0
20.9

19.8

IG

20.0

IGS
IS
S
SI

20.8
20.1

S
S
IG
IS
GS
IS
IS
IS
ISG
IG
I
IGS
IGS
I
I

25.8
20.1

9.8
16.1
17.6
19.8
16.1
11.6
19.9
19.3
17.6
20.8
23.3
4.0
2.4

9.9
16.7
17.3
15.7
16.9
7.3
19.0
*
16.0
18.9
19.5
1.5
5.3

SG
SGI
IG
**
S
GI
SI
I
GIS
I
I
ISG
ISG
I
I

Note: *specimens were broken along the interface during demolding. **the data was not recorded. I-failure occurred along interface; G-failure
occurred at the grout; and S-failure occurred at the substrate concrete.

41

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

3.2 Effects of Curing Temperature on Bond Strength
Fifteen grout products were selected for this test at three curing temperature levels (23°C, 30°C,
and 35°C). Table 2 summarizes all test results. The temperature change was found to havesubstantial impacts on the flexural bond strength. The 28-day flexural bond strength was
reduced with the increasing curing temperature possibly due to reduced flowability of fresh
grouts and poor quality of hydration products at higher curing temperature. Only a few products
showed the highest flexural bond strength at 30°C. Similarly, a high curing temperature
generally resulted in a reduced slant shear bond strength.
This can be again attributed to the
combined effect of the lowered flowability and a reduced quality of hydration products as thecuring temperature became higher. This also explain why the composite specimens were more
likely to fail at the interface at a high temperature level.
Table 2. Effects of temperature on the 28-day bond strength and failure plane of grout products
Product
ID

#1
#2
#3
#4
#5
#6
#7
#9
#10
#14
#19

23°C
MPa
FP
4.1
5.7
5.2
4.5
4.6
3.9
1.8
3.6
4.1
2.3

IS
S
IGS
I
S
G
GI
G
G
I
GS

Flexural bond test
30°C
MPa
FP
1.5
2.6
3.3
0.7
3.8
4.5
1.4
3.4
3.2
1.3

I
I
G
I
**
S
I
**
G
I
S
I
I

35°C
MPa
FP

23°C
MPa
FP

2.0
0.9
1.9
0.6
2.6
0.6
NA
1.6
2.8
0.2

19.1
22.3
20.2
10.4
26.0
22.2
10.0
18.6
19.8
16.1

I
I
G
I
I
I
NA
I
G
I
NA
G
I

Slant shear bond test
30°C
MPa
FP

MPa

35°C

FP

S
14.8
—
10.6
IG
IS
S
I
11.5
7.7
IGS
I
IG
11.8
14.3
I
19.8
IS
16.5
IG
IGS
21.0
—
17.4
IG
IGS
GS
I
17.2
9.8
GI
8.9
GIS
NA
NA
S
15.3
—
15.9
IG
IG
18.5
IG
11.6
I
IS
15.2
IS
15.3
IG
NA
IGS
ISG
1.8
4.5
NA
19.9
18.9
NA
#20
2.2
I
0.6
1.3
19.3
IG
17.4
IG
19.7
IG
#21
I
IG
I
I
2.9
1.8
2.2
17.6
9.4
16.7
#22
1.5
I
4.1
S
3.5
I
20.8
IGS
15.9
IGS
18.4
IG
#23
I
I
I
IG
IG
IGS
1.0
2.3
1.3
23.3
15.5
17.9
Note: *specimens were broken along the interface during demolding. **the data was not recorded. I-failure occurred along interface; G-failure
occurred at the grout; and S-failure occurred at the substrate concrete.

Steel fibers had distinct influences on the bond behaviors of composite specimens. Eight
products showed a reduced flexural bond strength because of 3% steel fiber additions. These
products mostly had high flexural bond strength (above 600 psi). This suggested that for grout
products already having high flexural bond strength, steel fiber addition would have negative
impacts on the bond strength. Conversely, six products showed an improved flexural bondstrength as 3% steel fibers were added. These products typically had relatively low flexuralbond strength (less than 600 psi), implying that adding steel fibers was beneficial in improving
the flexural bond strength of these products. The benefit of steel fiber addition was also reflectedby the failure plane. Without steel fibers, brittle failure typically occurred. When steel fibers
were added, they typically bridged the cracks, leading to tougher and more ductile failure.
Similar effects of adding steel fibers were observed for the slant shear bond test.
4. Acknowledgements
The author acknowledges the financial support from the Tennessee State Research and Planning(SPR) Program by the Tennessee Department of Transportation and the US Federal Highway
Administration. The author is also grateful for the assistance from the TDOT Study Advisory
Committee and from the lab manager and the students in SCCM of MTSU.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

5. References
ASTM C78-09, Standard Test Method for Flexural Strength of Concrete (Using Simple Beamwith Third-Point Loading), 4pp.
ASTM C882-13, Standard Test Method for Bond Strength of Epoxy-Resin Systems Used with
Concrete by Slant Shear, 4pp.

43

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Experimental Investigation of the Effect of Debonding on PSC Girders
Shambhavi Dube1* and Durgesh C. Rai2
1

: Doctoral Student, Dept. of Civil Engineering, Indian Institute of Technology Kanpur, Kanpur
208016, India; email: shamdube@iitk.ac.in2: Professor, Dept. of Civil Engineering, Indian Institute of Technology Kanpur, Kanpur 208016,
India; email: dcrai@iitk.ac.in*: corresponding author

Keywords: Pretensioned girder; partial debonding; shear strength; excessive debonding
Abstract: In pretensioned girders, the straight strands are debonded (sheathed) to follow theexternal moment profile, which also reduces unwanted stresses and cracks in the end region of
the girders. However, excessive debonding has detrimental effects on the shear strength of thegirder. In the past, some major bridge projects in India suffered significant web shear cracking
because of excessive debonding that was provided mainly based on loading requirements. To
experimentally investigate the influence of partial debonding on the shear strength of pretensioned concrete bridge girders, I-shaped half-scale fully bonded and partially debonded
girders have been designed, detailed, and fabricated in the laboratory at IIT Kanpur, India. Bothends of the girders are tested to compare the shear strength, stiffness, failure modes, and crack
patterns. The test results indicated that the ultimate strength of the girder reduces significantly indebonded girders compared to fully bonded girders. The ultimate failure in the fully bonded
girder was dominated by flexure shear crack in the bottom flange, whereas diagonal shear failurein the web was the primary failure mode for debonded girder.
1. Introduction
Debonding is performed in pretensioned bridge girders to follow the external bending moment
profile thereby reducing end forces This is achieved by breaking the bond between concrete andstrands either through hard debonding with a rigid tube or soft debonding with a split sheath.
Debonding is widely preferred over harping for its ease of manufacturing, making pretensioned
girders a cost-effective solution for the industry. However, excessive debonding can result in
reduced shear capacity and the appearance of diagonal cracks in the end region of girders which
has been a persistent challenge. The AASHTO LRFD currently limits the amount of partial
debonding to twenty-five percent of the total strand area within a pretensioned girder due to thedetrimental effects that excessive debonding can have on the shear performance of girders.
Nevertheless, other countries and some US states allow a higher percentage of debonding, up toseventy-five percent. The recent NCHRP Research Report 849 recommends going as high as
sixty percent debonding of the total strand area. The recent version of the Indian Road Congress
in its code (IRC 112-2020, Code of practice for concrete road bridges) permits debonding andlimits it to thirty-three percent through its special publication IRC: SP:71-2018. As the earlier
version of Indian code IRC 112-2011 had permitted debonding but was silent on the amount ofdebonding, some past bridge projects in India constructed before 2018 provided excessive
debonding (up to seventy percent) following the flexural requirement and experienced visible

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

distress in the form of inclined web cracking. Currently, there is no unanimity on the degree of
debonding among the various state of practice in the world. This experimental study is anattempt to improve current practices regarding the percentage of debonding followed worldwide.
2. Design and Fabrication of Specimen
Two pretensioned girders, fully bonded and 25% debonded, each 6 m in length and 800 mm indepth, were designed, detailed, fabricated, and tested for shear in the laboratory at IIT Kanpur,
India. Both identical ends (A and B) of each girder were tested, resulting in four sets of data as
shown in Fig. 1(a). Fig. 1(b) illustrates the cross-sectional details of the test specimen. In the
25% debonded girder, strands were debonded using a flexible tube to reduce prestress transfer
shown in Fig. 1(c). Both girders were cast and prestressed together in a 13 m long casting bed asshown in Figs. 1(d)-(f). To prevent cracking, a 28-day curing period was selected for transferring
the prestress into the concrete and a 28-day concrete strength of 60 MPa was achieved. The 7wire pretensioned strands made of 1860 MPa low relaxation steel having a diameter of 9.5 mm
were used. The experimental data indicated a Young's Modulus of around 230 GPa for all cases.
The 2-legged 6 mm dia. @ 150 mm c/c transverse stirrups of 6 mm dia. mild steel having a yieldstrength of 550 MPa, as obtained from the tensile test of the bar were used.

(a)

(b)

(c)
(d)
(e)
(f)
Fig. 1. Details of the fabrication (a) plan view; (b) cross-sectional view; (c) debonding of
strands; (d) casting bed; (e) close-up view of stirrups and formwork; (f) removal of formwork
3. Test Setup and Instrumentation
At each end of the girder, 20 strain gauges were placed in pretensioned strands and 12 wereplaced in transverse stirrups at various cross-sections, as illustrated in Figs. 2 (a)-(d). Four
additional strain gauges were installed at both ends of the debonded girder, at two sections: thepoint of debonding termination and a transfer length distance away from that point. A
pretensioned force of 60 kN (i.e., prestress of 1095 MPa) was applied to each of the 17 strands
symmetrically as shown in Figs. 2 (e) and (f) and measured using load cells at the dead end (seeFig. 2 (g)). Pretensioned strands were cut symmetrically in the cross-section using a gas cutter
from the center gap between the two girders such that pretension force transfers to both girders
simultaneously. Camber was measured for both girders using five LVDTs at L/6, L/3, L/2, 4L/6,
and 5L/6 from the girder end (where L is the length of girder) and strains were recorded. The
simply supported girder was mounted for testing on a Universal Testing Machine with a capacityof 2500 kN as shown in Fig. 3(a). The shear test was conducted on the girder with a span length

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Bridge Engineering Institute Conference 2023 (BEI-2023)
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of 2770 mm, with the actuator load applied 1125 mm from the right end. Five LVDTs wereplaced at various points along the girder to measure the vertical deflection, and the horizontal
movement was also recorded using another LVDT as shown in Fig. 3(b). Extensometer and
LVDT were placed diagonally within the shear span to record compression and extension, whichwere later used to calculate the shear strains.
(a)

(b)

(c)

Fully bonded strands
Debonded strands
Instrumented strands

(d)

(e)

(g)

(f)

Fig. 2. Details of the test specimen (a) sections in plan view; (b) detailed strand profile in eachcross-section; (c) application of narrow strain gauge; (d) strand and transverse strain gauges; (e)hydraulic jack for prestressing; (f) locked strands; (g) load cell to monitor applied prestress

(a)

(b)

Fig. 3. (a) Test girder; (b) test span and instrumentation (LVDTs and extensometer)
4. Results and Discussion
The applied load has been plotted with measured deflection under the applied load for both ends
of fully bonded and 25% debonded girders as shown in Fig. 4 (a). The experimental results havealso been compared with theoretical deflections up to the elastic limit and deflections fromABAQUS analysis for fully bonded case. The girders achieved their predicted capacity andexceeded the capacities calculated using AASHTO LRFD standards. The first crack appeared ata load of 11% and 16% more than the predicted capacities obtained from AASHTO LRFD for
fully bonded and 25% debonded girders, respectively. The maximum applied loads wereapproximately twice the load at which the first crack appeared (1211 kN and 1192 kN for fully46

Bridge Engineering Institute Conference 2023 (BEI-2023)
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bonded and 988 kN and 1068 kN for 25% debonded case; ends A and B respectively). The
results of the debonded girder tests indicated a reduction in shear strength. Prior to theoccurrence of the first crack, the stiffness of the debonded girder was comparable to that of afully bonded girder. However, once the crack appeared, the stiffness of the debonded girderdegraded more rapidly. Although the difference in stiffness was not significant at 25%
debonding, it is predicted to increase with higher debonding levels. The results clearly
demonstrate that debonding led to an increase in shear strain and a decrease in shear strength incomparison to the fully bonded case. An extensometer and LVDT were placed diagonally tomeasure compression and extension in the region between the support and load (shear span), andthe shear strain was calculated from this diagonal deformation. Figs. 4 (b) and (c) show the
relationship between applied shear and calculated shear strains for both ends of fully bonded and
25% debonded girders and compared them with theoretical elastic shear strain. In general, for aparticular value of applied shear, the shear strain was larger in debonded girder. As the girders
were tested up to failure, crack angles are discernible in both fully bonded and 25% debonded
girders as shown in Fig. 5. It is evident that the 25% debonded girder experienced more severe
cracking and breakdown in the lower flange compared to the fully bonded girder, which had ahigher level of prestress in the lower flange. Flexure shear cracks dominated the ultimate
behavior of the fully bonded girder, whereas, in debonded girder, extensive web shear cracking
governed the failure mode.

(a)

(b)

(c)

Fig. 4. (a) Applied load vs deflection plot; shear force vs deformation for: (b) fully bonded
girder; (c) 25% debonded girder

(a)
(b)
(c)
(d)
Fig. 5. Cracks and failure patterns: fully bonded girder (a) end A; (b) end B; 25% debonded
girder (c) end A; (d) end B
5. Conclusion
There was a notable reduction in the shear strength of 25% debonded girder and a rapid drop in
stiffness of debonded girder as compared to the fully bonded girder after the first crack.
Although the difference in stiffness was not significant at 25% debonding, it is predicted to47

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increase with higher debonding levels. The 25% debonded girder exhibited a wider and increased
number of diagonal cracks in the web compared to the fully bonded girder. To compensate forthe reduction in shear strength and stiffness, either more stirrups or an increased web width isrequired. An extensive experimental study is needed to address this issue.
6. References
AASHTO. 2017. LRFD Bridge Design Specifications, Eighth Edition, American Association ofState Highway and Transportation Officials, Washington D.C.
IRC 112, Indian Road Congress. 2020. Code of Practice for Concrete Road Bridges, IRC: 1122020, New Delhi, India.
IRC 112, Indian Road Congress. 2011. Code of Practice for Concrete Road Bridges, IRC: 1122011, New Delhi, India.
IRC, SP:71. 2018. Guidelines for design and construction of precast pre-tensioned girders forbridges, IRC: SP:71. 2018, New Delhi, India.
NCHRP 849 Research Report. 2017. Strand Debonding for Pretensioned Girders, National
Cooperative Highway Research Program, Washington, D.C.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Investigation of Nitrogen Oxide Removal of
Cementitious Materials Containing TiO2
Byoungsun Park1*, Suji Woo2, Young Cheol Choi3*, and Sung-Won Yoo4*
1: Department of Environmental Systems Engineering, Sejong Campus, Korea University,
Sejong-si 30019, South Korea; email: bspark0927@korea.ac.kr
2: Department of Civil and Environmental Engineering, Gachon University, Seongnam 13120,
South Korea
3: Department of Civil and Environmental Engineering, Gachon University, Seongnam 13120,
South Korea; email: zerofe@gachon.ac.kr
4: Department of Civil and Environmental Engineering, Gachon University, Seongnam 13120,
South Korea; email: imysw@gachon.ac.kr*: corresponding author

Keywords: Nitrogen oxide removal; Titanium oxide; Cementitious materials; ISO-22197-1;
photocatalytic reaction.
Abstract: Nitrogen oxide (NOx) is one of the primary contaminated matters in the air
environments. NOx is harmful to the human body and causes respiratory diseases. Therefore,various technologies to remove NOx are studied. It is known that NOx can be removed using aphotocatalytic reaction. Since the concrete structure has a large surface area and is exposed to theatmosphere, if a photocatalyst is mixed with the concrete so that the photocatalytic reaction canoccur on the concrete surface, NOx removal from the atmosphere can be efficiently performed.
This paper investigates NOx removal characteristics of cementitious materials containing
titanium oxide (TiO2). TiO2 is the most widely used photocatalyst because it is inexpensive andconvenient to use. TiO2 powder was added during the mixing process of cementitious materials
and was used in a ratio that partially replaces cement. Two types of cement composites (cement
pastes and foam composites) were used. NOx removal characteristics were investigated using themethod presented in ISO-22197-1. It was investigated that the NOx removal efficiency improved
as the TiO2 substitution rate increased. It was investigated that the NOx removal performance offoam composites was superior to that of cement pastes. In addition, the NOx removal efficiency
was found to be affected by the thickness of the specimen.
1. Introduction
Nitrogen oxide (NOx) is one of the main causes of air pollution and has harmful effects on thehuman body (Seo and Yun, 2017). In addition, NOx is one of the precursors of particulate matter.
Therefore, various studies are being conducted to remove NOx in the air (Chen at al., 2021). Oneof them is to use a photocatalytic reaction (Yu and Brouwers, 2009). In the photocatalytic
reaction, NOx is oxidized to NO3- using ultraviolet (UV) light, and NO3- is removed by water (Yuand Brouwers, 2009). TiO2 is the most widely used photocatalyst due to its economic feasibility
and ease of use (Yu at al. 2006). Since the photocatalytic reaction requires UV light, it is49

Bridge Engineering Institute Conference 2023 (BEI-2023)
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important to expose the photocatalyst to the outside. The concrete structure has a high surface
area and is suitable as a site for photocatalytic reaction because it is built on the outside. Since
the photocatalytic reaction appears only at the location where UV light is supplied, it is necessary
to increase the amount of photocatalyst exposed on the concrete surface.
In this study, the NOx removal performance of TiO2-added cementitious composites wasinvestigated. An experiment was conducted using foam composite and cement paste toinvestigate the effect of concrete surface area on NOx removal performance.
2. Experiments
In this study, ordinary Portland cement (OPC) was used as binder. The density and Blainfineness of OPC were 3.1 g/cm3 and 3,260 cm2/g, respectively. The average particle size of OPC
was 5.9 μm. TiO2 in powder form was added to cementitious materials to impart photocatalytic
reactivity. The ratio of anatase and rutile constituting TiO2 was 8:2. Table 1 shows the mixture
proportions of specimens. Cement pastes and foam composites were used to investigate theeffect of porosity of cementitious materials on NOx removal performance. Water to cement ratio
is fixed as 0.3. The foaming ratio of foam composites was a fix at 63.6%. OPC was substituted as
TiO2 powder in the ratio of 3% by volume. The specimens were fabricated in the size of
200×50×20mm. The specimens were cured in the chamber at 23 ± 2 °C in temperature and 60 ±10 % in relative humidity.
Table 1. Mixture proportions of specimens
W/C
OPC [kg/m3]
Plain
500
PlainT3
485
0.3
Foam
500
FoamT3
475

TiO2 [kg/m3]
21.9
21.9

Foaming ratio (%)
63.6

To evaluate the NOx removal performance of cementitious materials, a test equipment withreference to ISO-22197-1 was prepared. The test equipment consists of NOx concentration
analyzer, photocatalytic reaction chamber, UV lamp, NO gas and air supply equipment. Test
specimens fabricated according to Table 1 were pretreated to remove contaminants on thesurface using a UV lamp and a drying chamber. The pre-treated specimens were installed in thechamber. 1 ppm of NO gas was supplied at a rate of 3 L/min. UV light was irradiated for 3 hours
from 2 hours after the specimen was placed. The experiment was continued for 1 hour after
turning off the UV light.
3. Results & Discussion
Fig. 1 shows NOx concentration obtained from experiments using cement paste and foamcomposite. In the experiment using Plain, the NOx concentration reached about 0.95 ppm after
supplying NO gas. On the other hand, in the experiment using foam, the NOx concentration
reached 0.95ppm, then decreased, and then recovered after 2 hours. Since the foam has arelatively high surface area compared to the plain, it is judged that NOx is adsorbed. Due to thehigh surface area of the foam, the incorporation of TiO2 seems to increase the efficiency of thephotocatalytic reaction. In both specimens, NOx concentration was constant regardless of UV
light. This result means that NOx removal reaction does not appear without TiO2.

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Fig. 1. NOx concentration obtained from experiments using cement paste and foam composite
Fig. 2 shows NOx concentration obtained from experiments using specimens incorporating TiO2
powder. Fig. Unlike 1, NOx decreased after irradiation with UV light. This is due to thephotocatalytic reaction of TiO2. Looking at the graph, the NOx concentration of the two
specimens was the same immediately after irradiation with UV light. As the UV light irradiation
time increased, the NOx concentration of PlainT3 increased compared to that of FoamT3. Due tothe high surface area of Foam, more TiO2 powder was exposed to the surface, which seemed toincrease the efficiency of the photocatalytic reaction.

Fig. 2. NOx concentration obtained from experiments using specimens incorporating TiO2
powder
4. Conclusion
In this study, the NOx removal performance of TiO2-added foam composites and cement pastes
was investigated. Foam composites showed increased adsorption of NOx compared to cement51

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

pastes due to their high surface area. When TiO2 was added, the NOx removal performance offoam composites increased compared to that of cement pastes. It can be seen that the NOxremoval performance can be improved by exposing more TiO2 powder to the surface.
5. References
Seo, D. and Yun, T.S. 2017. NOx removal rate of photocatalytic cementitious materials with
TiO2 in wet condition, Building and environment, 112, 233-240.
Chen, R., Zhang, T., Guo, Y., Wang, J., Wei, J., and Yu, Q. 2021. Recent advances insimultaneous removal of SO2 and NOx from exhaust gases: Removal process, mechanism andkinetics, Chemical Engineering Journal, 420, 127588.
Yu, Q. and Brouwers, H.J.H. 2009, Indoor air purification using heterogeneous photocatalyticoxidation. Part I: Experimental study, Applied Catalysis B Environmental, 92, 454–461.
Yu, K.-P., Lee, G.W.M., Huang, W.-M., Wu, C., and Yang, S. 2006. The correlation between
photocatalytic oxidation performance and chemical/physical properties of indoor volatile organic
compounds, Atmospheric Environment, 40(2) 375-385.
6. Acknowledgement
This work was supported by the National Research Foundation of Korea (NRF) grant funded by
the Korea government (MSIT) (NRF-2020R1A2C2008926). This work was also supported by
the Korea Conformity Laboratories (PUR04202302280031).

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

The Diagnosis Method for Neutralization of Concrete Using Microorganisms
China Kuratomi1*, Atsushi Teramoto2, So Fujiyoshi3, and Fumito Maruyama4
1*: Hiroshima University, Higashi-Hiroshima, Japan; email: d231038@hiroshima-u.ac.jp: Hiroshima University, Higashi-Hiroshima, Japan; email: a-teramoto@hiroshima-u.ac.jp
3: Hiroshima University, Higashi-Hiroshima, Japan; email: sofu62@hiroshima-u.ac.jp
4: Hiroshima University, Higashi-Hiroshima, Japan; email: fumito@hiroshima-u.ac.jp
2*: corresponding author

Keywords: neutralization; microbiome; DNA extraction; water repellency
Abstract: The risk of rebar corrosion in concrete piers is increased by pH neutralization due tocarbonation. The progress of neutralization inside concrete is difficult to estimate through wholly
nondestructive methods. In this study, we examine a new possible diagnostic method forneutralization by clarifying how the biological community in concrete are affected by (1)environmental factors causing neutralization and (2) the concrete after neutralization has occurred.
First, specimens with different surface pore structures were prepared by changing the age of thedemolding. Some of the specimens were coated with water-repellent or hydrophilic paint.
Thereafter, the specimens were periodically moistened via two different supply methods: runningwater and spraying in accelerated carbonation period. After the accelerated carbonation test,
microbial samples were obtained from the surface and interior of the specimens to investigate thedifferences in microbial community characteristics. The results indicated that the pore structure of
the surface layer was the most significant factor affecting the rate of concrete neutralization.
Although the coating had the effect of delaying the age at which neutralization began, it did not
significantly affect the rate of subsequent neutralization progression, regardless of whether they
were hydrophilic or hydrophobic materials. To quantitatively evaluate the number ofmicroorganisms present at each location, DNA concentrations were compared. Moremicroorganisms were found to be present in the surface layer than in the interior, and the DNA
concentration in the surface layer was greater for concrete with a shorter demolded age and whenwater was supplied by running water, the DNA concentration in the surface layer decreased in thisexperiment. This indicates the effect of washing off microorganisms in the surface layer was
greater than the effect of increasing the moisture content of the specimens. In addition, for samples
taken from surface swabs, the results suggest that the Shannon diversity index is not dependent on
the pore structure of the concrete, but rather on the type of surface coating.
1. Introduction
Microorganisms are found worldwide, and certain microorganisms are thought to inhabit concrete.
Previous studies have shown that microorganism growth is characterized by the carbon dioxide
concentration (Balzani et al, 2022), temperature and humidity (Hoeksma et al, 2015), and pH(Preiss et al, 2015) of the surrounding environment. Therefore, microorganisms inhabiting
concrete might be characterized by the properties of the concrete and environment. External factors

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influencing the progress of concrete neutralization include carbon dioxide, ambient humidity andtemperature, and rainwater on the structure. In addition, the pH of the solution in the pores andpore size of the neutralized concrete changes. Therefore, if the properties of neutralized concrete
can be related to the microbial habitat, a new diagnostic method could possibly be developed.
In this study, the effects of external factors causing neutralization and concrete properties after
neutralization on the microorganisms in concrete were clarified. Moreover, the applicability of a
new diagnostic method developed for neutralization is discussed. Furthermore, the effects ofconcrete water supply, pore structure, and water repellency of the surface layer are described.
2. Experiment2.1. Test Specimen
Concrete specimens with a water cement ratio of 53% and internal dimensions of 200×130×75
mm were cast using square styrol cases. Various concrete specimens were fabricated to assess the
progression and velocity of neutralization using the following parameters: coarse and dense pore
structure, water repellency of the specimen surface, and moisture supply method.
The specimens cured for 28 days were fabricated with and without hydrophilic and hydrophobic
coatings (hereafter referred to as hydrophilic, hydrophobic, and 3w, respectively). In addition, thewater repellency of the surface layer was differentiated.
2.2. Accelerated Neutralization Tests
Accelerated neutralization tests were conducted on the specimens described in Section 2.1. To
limit the carbon dioxide infiltration path, the casting surface and four sides of the specimen weresealed with aluminum tape. Accelerated neutralization tests were conducted for up to 26 weeks at
20 °C with a relative humidity of 60% and carbon dioxide concentration of 5%, in accordance with
JIS A 1153. Neutralization depth measurements and water supply were conducted at accelerated
ages of 1, 4, 8, 13, and 26 weeks. The specimens for each parameter were watered with either aconstant amount of running water applied to the test surface (hereafter referred to as "runningwater") or sprayed with a sprayer to form a mist (hereafter referred to as "spraying"). For therunning water, a water reservoir was installed above the specimen with a partition raised to a fixed
position to allow a fixed amount of water to flow at an approximately constant rate. For thespraying, the amount of water supplied was controlled by the amount of water emitted from thesprayer and number of times sprayed. Specimens not subject to a water supply were alsoestablished to investigate the degree of neutralization enhancement with and without a watersupply.
2.3. Sampling
Samples for microbiological analysis were collected from the surface of the neutralized concrete
specimens by swabbing and from 10 mm inside by breaking the specimens with a hammer. The
26-week measurements of all samples were used for the accelerated neutralization tests. Swabbing
was applied twice to the same surface area of approximately 100×50 mm. The swab sticks andcrushed samples were stored at -80°C to inactivate the microorganisms and maintain the microflora
at the time of the DNA extraction procedure (Section 2.4).
2.4. DNA Concentration Measurement
DNA extraction was performed using the DNeasy® PowerBiofilm® Kit (QIAGEN) according toa developed protocol by the authors. The DNA concentration was measured by determining the54

Bridge Engineering Institute Conference 2023 (BEI-2023)
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3.2. DNA Concentration
The average swab area for all specimens was 5256 mm²,
and the average internal sample mass was 0.0649 g. The
preliminary experiment revealed that the mass of theadhered material collected from a 150×150 mm concrete
wall was 0.0239 g. The swab area in this experimentacquired approximately 0.0004 to 0.0040 g of adheredmaterial. Based on this acquired amount, the DNA
concentration, extracted from 1 g of the sample collectedby the swab method, was approximately 15 to 1300 times
greater than that of the internal sample. In comparison withprevious studies (Kuratomi et al, 2022), this order ofmagnitude is reasonable and reconfirms thatmicroorganisms are concentrated in the surface layer ofconcrete rather than the interior.
The DNA concentrations per swab area and per gram ofinternal sample were calculated by dividing the DNA
concentration extracted in this experiment by mass of theswabbing sample and the internal sample used forextraction, respectively. These DNA concentrations areillustrated in Figs. 5 and 6. The DNA concentration
detected via the swabbing method tended to be affected by
the water supply method, with higher DNA concentrations
extract in order: no water supply, spraying, running water
extracting. This trend was particularly noticeable for

55

Neutralization depth (mm)running70

A : 11.141

spraying

60
50
40

A : 10.732

30
20

running-approximation

100spraying-approximation
0

10

20

30

Weeks of measurement

Fig. 1. The 1d specimen
neutralization depth andneutralization rate coefficient A
Neutralization depth (mm)80
running
3w-流水70

spraying
3w-噴霧

60

running-approximation
3w-流水近似50

spraying-approximation
3w-噴霧近似40
30

A : 4.906

20
100

A : 3.856
0

10

20

30

Weeks of measurement

Fig. 2. The 3w specimen
neutralization depth andneutralization rate coefficient A
80
親水-流水
running70

Neutralization depth (mm)3. Results and Discussion
3.1. Accelerated Neutralization Depth
The neutralization depth, least-squares approximation,neutralization rate coefficient, and neutralization
resistance of each specimen were calculated, as shown in
Figs.1-4. The neutralization progression of the 1-day
specimens at 26 weeks was approximately three times
greater than that of the 28-day specimens. Demolding at a
young age were deemed to result in a coarse pore structure
thus facilitating carbon dioxide infiltration. Conversely, nodifference in the neutralization rate coefficient owing todifferences in the water supply method or water repellencyof the surface were observed.

80

spraying
親水-噴霧60

running-approximation
親水-流水近似50

spraying-approximation
親水-噴霧近似40
30

R : 1.694

20
100

R : 2.083
0

5

10

15

20

25

30

Weeks of measurement

Fig. 3. Hydrophilic specimen
neutralization depth andneutralization resistance R
80

Neutralization depth (mm)fluorescence using a trace spectrophotometer (DS-11FX,
Denovix). The resulting DNA solution was subjected to a
2-step tailed PCR method to create a library. Subsequently,
the sequencing was performed using the MiSeq system and
MiSeq Reagent Kit v3 (Illumina) at 2 × 300 bp.

running
疎水-流水70

spraying
疎水-噴霧60

running-approximation
疎水-流水近似50

spraying-approximation
疎水-噴霧近似40
30

R : 2.055

20
100

R : 2.036
0

10

20

Weeks of measurement

30

Fig. 4. Hydrophobic specimen
neutralization depth andneutralization resistance R

Concentration of DNA

spraying

without water supply

1

0.1

0.01

1d

3w

hydrophilic
親水
Test specimenhydrophobic疎水

DNA concentration per gramof swabbing sample
running

10
Concentration of DNA per gram of

4. Conclusion
In this study, accelerated neutralization tests wereconducted with different pore structures and surface
layer water repellency. And microorganisms weredetected on the surface and from the interior of theconcrete. As a result, DNA was extracted from surface
layer at concentrations several hundred times higher
than the samples taken from the interior concrete.
Microorganisms in concrete were found to besusceptible to the influences of external environmental
factors and the materials property of surface layer. The
above results may contribute to a significant
simplification of bridge maintenance, which iscurrently carried out at great cost.

running

10

Fig. 5.
internal concrete sample（ng/µL/g）

specimens without water repellency. For the interior
samples, the lowest DNA concentration was extracted
from the sprayed specimens in the test system withouta water-repellent coating, whereas the DNA
concentration of the specimens without water supply
tended to be lower in the test system with a waterrepellent coating. These results indicate thatmicroorganisms in concrete can be affected by watersupply conditions and surface water repellency.
Fig. 7 shows the Shannon diversity index calculated
from the obtained sequences data. The figure showsthat the diversity of the microbial community in thesurface layer of the hydrophilic paint film is higher
than others, suggesting that the type of paint film may
also affect the microbial community.

per gram of swabbing sample（ng/µL/g）

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

spraying

without water supply

1

0.1

0.01

1d

3w

親水
hydrophilic

疎水hydrophobic

Test specimen

Fig. 6. DNA concentration per gramof internal concrete sample

Fig.7 Shannon diversity of
microbiome on surface of specimens

5. References
Balzani et al. 2022. CO2 biogeochemical investigation and microbial characterization of res wood
ant mounds in a Southern Europe montane forest, Soil Biology and Biochem., 166, 1-10
Hoeksma et al. 2015. Effect of temperature and relative humidity on the survival of airborne
bacteria, Livestock Research Report, 864, 1-27
Kuratomi et al. 2022. The basic study of microbial detection method for concrete, Proceedings of
the Japan Concrete Institute, 44, 1300-1305
Preiss et al. 2015. Alkaliphilic bacteria with impact on industrial applications, concepts of earlylife forms, and bioenergetics of ATP synthesis, Frontiers in Bioeng. and Biotech., 75(3), 1-16

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6. Acknowledgments
This work was supported by JST FOREST Program, Grant Number JPMJFR215N and the Fundfor the Promotion of Joint International Research (Fostering Joint International Research (B)) of
KAKENHI 22KK0064. The hydrophobic materials were provided by Dr. Tsujino (Shimizu Corp.)
and Mr. Azuma (Toyo Aluminium K. K.)

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Bridge Engineering Institute Conference 2023 (BEI-2023)
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Research on Fatigue Properties of Basalt Fiber
Cement Stabilized Macadam
Wei Li1, Peifeng Cheng1, Wenmei Zhao1, and Zhanming Zhang1*
1: College of Civil Engineering and Transportation, Northeast Forestry University, Harbin
150040, China; email: alice12383@126.com*: corresponding author
Keywords: Cement stabilized aggregates; Basalt fiber; Fiber dosing; Stress level; Fatigue
performance
Abstract: In order to study the fatigue performance of basalt fiber-doped cement-stabilized
macadam, three-point small beam bending fatigue tests were conducted for cement-stabilized
macadam with different stress levels and basalt fiber doping. Based on the two-parameter
Weibull distribution theory, the fatigue test results were analyzed and the fatigue equation wasestablished. The mechanism of fatigue performance of basalt fiber reinforced cement stabilized
macadam is explained by combining with micro-morphological observation method. The results
show that basalt fibers are uniformly distributed in cement-stabilized macadam to form a
network structure, so that the material is uniformly stressed and load concentration is avoided,
and the reinforcing effect delays the expansion of fatigue micro-cracks. The fatigue life of
cement-stabilized macadam can be increased by 2-5 times by adding basalt fibers at different
stress levels. At a stress ratio of 0.6, the fatigue life of cement-stabilized macadam was increasedby more than 5 times. In addition, the constructed fatigue prediction equation can be used to
predict the fatigue life of cement-stabilized macadam at different load levels, which can be used
as a reference for engineering applications.
1. Introduction
Cement-stabilized crushed stone materials are widely utilized in transportation engineering
construction due to their notable characteristics, including high overall integrity and resistance todeformation. However, the combined effects of factors such as load, temperature, and humidity
can compromise the tensile strength and deformability of cement-stabilized crushed stone,
leading to the formation of cracks under fatigue conditions and consequent reduction in loadbearing capacity. Therefore, the prevention and mitigation of crack formation play a crucial rolein enhancing the fatigue life and durability of cement-stabilized crushed stone [1]. The
incorporation of fibers in cement-stabilized crushed stone materials has been found to improvetheir mechanical properties and resistance to fatigue cracking [2]. Basalt has been successfully
employed in engineering applications to enhance the performance of asphalt mixtures [3],mitigate deterioration under aging conditions, and improve the mechanical properties of cement
concrete and mortar[4].In this research, the focus is on studying and applying the fatigue
performance of cement-stabilized crushed stone materials reinforced with basalt fibers. The
fatigue life is evaluated by conducting tests at different stress ratios and fiber contents. The
analysis of fatigue test results and the development of fatigue equations are performed based on58Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

the two-parameter Weibull theory. This study aims to provide valuable insights and guidance forinvestigating and utilizing the fatigue resistance of basalt fiber-reinforced cement-stabilized
crushed stone materials.
2. Test Materials and Methods
2.1 Material
The experimental materials primarily consist of three main components: cement, basalt fibers,
and aggregates. The cement material selected is ordinary Portland cement with a strength gradeof 42.5, produced by a chemical plant in Heilongjiang Province, China. The cement content is
4.5%, meeting all the specifications required by the "Test Code for Cement and Cement
Concrete in Highway Engineering" (JTG 3420-2020). The basalt fibers used are speciallyprovided by a composite materials company in Zhejiang Province. These fibers are in the form of
bundled cement composite short-cut strands, specifically designed for this study. The main
performance indicators of the basalt fibers can be found in Table 1. The aggregates used in theexperiment were sourced from a limestone quarry at a highway construction site in Heilongjiang
Province, China. The primary component of the aggregates is limestone ore.
Table. 1. Main technical indicators of cement
Standard
consistency(%)

Initialsetting time
(min)

Final
setting time(min)

Stability(mm)

3d Flexuralstrength
(MPa)

27.5

325

412

qualified

5.6

3dCompressive
strength
(MPa)
22.4

2.2 Fatigue test method
The fatigue tests were performed using a UTM-30 universal testing machine. The loading
method selected was a stress-controlled mode, which closely mimics the actual working
conditions and ensures good reproducibility. The tests were conducted at a temperature of 15°C
with a loading frequency of 10Hz. A half-sine waveform was employed as the load waveform,
representing the ratio between the minimum and maximum cyclic stresses. The fiber dosages
chosen were 0‰, 0.4‰, 0.6‰, and 0.8‰, while the stress levels for the fatigue tests were set at
0.6, 0.7, 0.75, and 0.8.For the fatigue tests on cement-stabilized crushed stone incorporatingbasalt fiber, standard small beam specimens with dimensions of 100mm × 100mm × 400mm
were prepared. These specimens were subjected to a curing period of 90 days before testing. To
minimize experimental errors, each group of samples underwent five parallel tests.
3. Results and Discussion
3.1 Fatigue Test Results
The results of the fatigue tests on cement-stabilized crushed stone specimens with different
basalt fiber dosages are presented in Table 2, with the flexural strength based on various basalt
fiber dosages as the reference. Within the stress range of 0.6 to 0.8, under the same gradation andcement content, the addition of basalt fibers effectively enhances the fatigue life of the cementstabilized crushed stone specimens. The fatigue life of basalt fiber-reinforced cement-stabilized
crushed stone is significantly improved, with an enhancement ratio ranging from 2 to 5 times.
Under a stress ratio of 0.6, it can achieve a maximum increase of 5.61 times compared toordinary cement-stabilized crushed stone.

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Table. 2. Fatigue test results of basalt fiber cement stabilized gravel
Average fatigue life (times)

Stresslevel

Fiber content 0‰

Fiber content 0.4‰

Fiber content 0.6‰

Fiber content 0.8‰

0.80

7810

17921

25220

34217

0.75

25163

58746

84422

93531

0.70

1743824233877191859149150.60

89125827561244325547

5000000*

Note: * indicates specimens that reached 5,000,000 cycles without failure.
To examine the relationship between the fatigue life of cement-stabilized crushed stone withbasalt fiber and the dosage of basalt fiber, a curve fitting analysis was conducted. The basalt fiber
dosage was selected as the independent variable (x-axis), while the logarithm of the averagefatigue life was plotted as the dependent variable (y-axis). This analysis yielded the relationshipbetween basalt fiber dosage and fatigue life at different stress levels, as depicted in Fig.1.

(a) Stress ratio of 0.6(b) Stress ratio of 0.7(c) Stress ratio of 0.75(d) Stress ratio of 0.8

Fig.1. Relationship between basalt fiber doping and lgN for different stress ratio levels
lgN = -0.8089x2 + 1.5970x + 5.9473
lgN = -0.2864x2 + 1.1495x + 5.2374
lgN = -0.6007x2 + 1.2062x + 4.3982
lgN = -0.2448x2 + 0.9971x + 3.8928

60

(1)
(2)
(3)
(4)

Bridge Engineering Institute Conference 2023 (BEI-2023)
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From Fig.1, it is evident that the logarithm of the average fatigue life is well-fitted by a quadratic
parabolic relationship with the dosage of basalt fibers. For stress levels of 0.6, 0.7, 0.75, and 0.8,
the fitted quadratic parabolic equations describing the relationship between basalt fiber dosage
and fatigue life are represented by Equations 1 to 4, respectively. In these equations, x denotes
the dosage of basalt fibers, and N represents the number of fatigue cycles. It can be observed thatthe fatigue life values at different stress levels exhibit a monotonically increasing trend with anincrease in basalt fiber dosage. However, as the stress level gradually increases, the fatigue lifeof basalt fiber-reinforced cement-stabilized crushed stone decreases.
3.2 Build the fatigue equation
In fatigue testing, it is common practice to utilize the S-N (Stress level-Fatigue life) curve todepict the experimental results, which aids in predicting the fatigue life of basalt fiber-reinforced
cement-stabilized crushed stone under various stress levels. This approach holds significant
engineering significance as it provides valuable guidance. An exponential function-type singlelogarithm fatigue equation is chosen, as represented by Equation 5.
（5）
After taking the logarithm of both sides, Equation 5 can be transformed into Equation 6：（6）
Where：A=lg C/Qlge，B=1/ Qlge，where Q and C are arbitrary constants.
3.3 Mechanism of anti-fatigue performance of basalt fiber reinforced cement stabilized
macadam
In order to investigate the mechanism behind the enhanced fatigue resistance of basalt fiberreinforced cement-stabilized crushed stone, the bonding between basalt fibers and thecementitious matrix, as well as the microstructural changes of basalt fibers under fatigue failure
conditions, were examined. Microscopic images of the basalt fiber-reinforced cement-stabilized
crushed stone in its normal state and at the fractured interface were captured using a JSM-7500F
scanning electron microscope, as shown in Fig.2. From Fig.2(a), it can be observed that thebasalt fibers are embedded within the cement-stabilized crushed stone, providing a structure that
disperses stress and promotes a more even stress distribution, thereby preventing stressconcentration. When the cement-stabilized crushed stone is subjected to repeated loading or
temperature fluctuations leading to fatigue cracks, the basalt fibers prevent further crack
propagation and act as reinforcement by bridging the fractured sections. Fig.3(b) reveals a dense
and well-bonded interface between the fibers and the cement-stabilized crushed stone, without
any signs of loose fibers being pulled out. Additionally, evidence of fibers being pulled andfractured within the mixture can be observed, indicating the fibers' effective crack-arresting
ability and their role in enhancing the fatigue crack resistance of the cement-stabilized crushedstone material. During the micro-damage stage, the expansion of cracks encounters a dense
network of randomly distributed fibers within the matrix, which effectively impedes fatigue
damage, delays unstable crack propagation, and mitigates failure occurrence.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023(a)Normal section(b)Destruction section

Fig.2. Microstructure of basalt fiber cement stabilized gravel
4. Conclusions
1. The addition of basalt fibers significantly enhances the fatigue life of cement-stabilized
crushed stone. The inclusion of basalt fibers can increase the fatigue life of cement-stabilized
crushed stone material by a substantial margin, ranging from 2 to 5 times. Under a stress ratio of
0.6, the incorporation of basalt fibers can even boost the fatigue life by up to 5.61 times.
2. Through regression analysis based on the Weibull distribution theory, a well-correlated fatigue
equation has been developed to predict the fatigue life of cement-stabilized crushed stonematerial. This equation provides a reliable tool for estimating the fatigue performance of suchmaterials.
3. Microscopic testing has revealed the beneficial effects of incorporating basalt fibers in the
cement matrix. These fibers help disperse stress, prevent stress concentration, and effectively
inhibit crack propagation within the mixture. As a result, the resistance of cement-stabilized
crushed stone material to fatigue-induced cracking is significantly enhanced.
5. References[1]
Baghini M S, Ismail A, Karim M R, et al. Effect of styrene-butadiene copolymer latex on
properties and durability of road base stabilized with Portland cement additive [J]. Construction
And Building Materials, 2014, 68: 740-9.
[2]
Ma Y H, Gu J Y, Li Y, Et Al. The Bending Fatigue Performance of Cement-Stabilized
Aggregate Reinforced with Polypropylene Filament Fiber [J]. Construction And Building
Materials, 2015, 83: 230-6.
[3]
Pirmohammad S, Amani B, Shokorlou Y M. The Effect of Basalt Fibres on Fracture
Toughness of Asphalt Mixture [J]. Fatigue & Fracture of Engineering Materials & Structures,
2020, 43(7): 1446-60.
[4]
Sarkar A, Hajihosseini M. The effect of basalt fibre on the mechanical performance ofconcrete pavement [J]. Road Materials And Pavement Design, 2020, 21(6): 1726-37.

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Wind, Vibration, and Aerodynamics

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Analysis of Dynamic Characteristics of Semi-Submersible Deep-Water
Floating Foundation under Wind-Wave Coupling Actions
Honggang, Xu1, Haiquan, Jing2*, and Xuhui, He3
1: School of Civil Engineering, Central South University, Changsha, China; email:xuhonggang1992@126.com
2: School of Civil Engineering, Central South University, Changsha, China; email:hq.jing@csu.edu.cn
3: School of Civil Engineering, Central South University, Changsha, China: email:xuhuihe@csu.edu.cn*: corresponding author

Keywords: Deep-water floating foundation; Wind-wave coupling actions; Taut mooring cables;
Dynamic response analysis
Abstract: The deep-water floating bridge is a novel concept aimed at solving the challenge of
constructing traditional bridges in wide and deep fjords. Compared to conventional bridges, thedeep-water floating bridge has significant advantages, including lower construction costs andbetter environmental protection. This paper proposes a semi-submersible deep-water floating
foundation and conducted an analysis of the coupled dynamic characteristics of the semisubmersible deep-water floating foundation under the coupling actions of wind and waves byusing ANSYS-AQWA. The research results indicate that wind loads have a controlling effect on
the displacement response of the structure, while wave loads have a controlling effect on theacceleration response of the structure.
1. Introduction
As a novel form of bridge, the deep-water floating bridge can effectively solve the technicalchallenges of building traditional bridges in wide sea areas with high water depths. However,compared with traditional bridges, deep-water floating bridges are more susceptible toenvironmental loads. Therefore, accurately assessing their structural dynamic response is of great
importance.
Many scholars have conducted extensive research on the structural form and dynamic
response of deep-water floating bridges. Tang M et al. studied the hydrodynamic characteristics
of a certain pontoon-type floating bridge in deep water under the influence of different
foundation shapes and structural parameters based on the hydrodynamic calculation software
AQWA. The results showed that a reasonable size of the windward face and interface shape cansignificantly reduce the wave force on the structure. Moreover, it was found that thehydrodynamic characteristics and economy of a semi-submersible foundation are significantly
better than those of a fully submerged one (Tang et al. 2021). Wei K et al. used a tension leg
foundation deep-water floating bridge as an example and analyzed the influence of tension leg
parameters on the foundation stiffness and dynamic response of the floating bridge by64

Bridge Engineering Institute Conference 2023 (BEI-2023)
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calculating the pulsating wind load and wave load using the finite element analysis method. The
research results showed that increasing the immersion depth of the foundation and the inclination
angle of the tension leg can effectively reduce the structural dynamic response (Wei et al. 2022).
Currently, research on deep-water floating bridges is still in the theoretical stage, and there
are no engineering examples of such bridges worldwide. This article proposes a semisubmersible deep-water floating foundation. Different marine environments and load
combinations are studied to investigate their effects on the structural dynamic characteristics.
2. Dynamic response analysis

Fig. 1. Illustrated diagram of deep-water floating foundation
The proposed semi-submerged deep-water floating foundation in this paper consists of several
key components, including the upper support platform, columns, lower floating box, andmooring system. The upper support platform supports the upper bridge structure and transfers thegravity load to the lower foundation. Its dimensions are determined by the bridge tower andlower floating box and it adopts a hollow box design. The columns connect the upper support
platform and lower floating box and can also provide some buoyancy. The lower floating box
mainly provides the required buoyancy for the floating foundation, and also has a ballast tankthat provides restoring torque to maintain the structure vertically under external loads. It can also
be used to adjust the overall center of gravity of the structure to maintain stability. The mooring
system connects the floating foundation to the seabed and limits the overall displacement of thestructure.
In order to investigate the role played by wind and wave loads in the coupling effect ofwind and waves under different marine environments, this study examines the dynamic
responses of structures under different load combinations. Specifically, this study investigates
three load combinations: wind load only, wave load only, and wind-wave coupling. The wave
parameters are determined based on the wind-wave deterministic relationship proposed by China(JTS 145-2015), under the given conditions of wind speed at 10 meters above sea level. The
wind and wave parameters of each case are shown in Table.1.
Table 1. Characteristic parameters of wind and waves

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U10 (m/s)
Hs (m)Ts (s)
Case 1
10
0. 80
3.28
Case 220
1.96
4.75
Case 330
3.26
5.83
Case 440
4.58
6.64
Case 550
5.82
7.23
Figure 2 illustrates the displacement and acceleration responses of the structure under
different loading combinations as the wind speed U10 increases from 10 m/s to 50 m/s. It isobserved that both the displacement and acceleration responses of the structure increase with theincrease of wind speed. The displacement and acceleration responses of the structure underwind-wave coupling are greater than those under wind load or wave load alone, and thestructural response under coupling effects is not simply the superposition of the responses underindividual loading conditions. When U10=10 m/s, the displacement response of the structure
under wind load alone is greater than that under wave load alone, but the two loading conditions
cause similar displacement responses. As wind speed increases, the displacement response of thestructure under wind load alone becomes significantly greater than that under wave load alone,indicating that wind load plays a dominant role in controlling the displacement response of thestructure. On the contrary, the acceleration response of the structure under wave load alone ismuch greater than that under wind load alone, indicating that wave load plays a dominant role incontrolling the acceleration response of the structure.

(a)

(b)

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(c)

(d)

(e)
Fig. 2. Comparisons of structural dynamic response under various loading combinations:(a)U10=10m/s; (b) U10=20m/s; (c) U10=30m/s; (d) U10=40m/s; (e) U10=50m/s
3. Conclusion
This article proposes a semi-submersible deep-water floating foundation and analyzes thecoupled dynamic characteristics of the semi-submersible deep-water floating foundation under
the coupling effect of wind and waves. The role of wind and wave loads in the coupling effect ofwind and waves under different marine environments is studied. The research results indicatethat the coupling effect of wind and waves is not simply the superposition of wind load and wave
load. Wind load plays a controlling role in the displacement response of the structure, while
wave load plays a controlling role in the acceleration response of the structure.
4. ReferencesWei, K. Zhang, F. Liao, X. and Qin, S.Q. 2023. Effect of tension leg foundation stiffness on
dynamic responses of a long-span floating suspension bridge under wind and wave loads, 55(6),47-62.

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Tang, M. Xiang, S and Cheng, B. 2021. Hydrodynamic property analysis of pontoon pile cap of
deep-water floating bridge, 49(6), 78-84.
JTS 145-2015. Code of Hydrology for Harbour and Waterway.

68

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Field Measurement Study on Wind Characteristics
at the Bridge Site of Plateau Deep-Cut Canyon
Xinghui Kang1, Yunfeng Zou2*, Xuhui He3, and Longan Li4
1: School of Civil Engineering, Central South University, Changsha, China; email:
xinghuikang@csu.edu.cn
2: School of Civil Engineering, Central South University, Changsha, China; email:yunfengzou@csu.edu.cn
3: School of Civil Engineering, Central South University, Changsha, China; email:xuhuihe@csu.edu.cn
4: China Railway Major Bridge Reconnaissance & Design Institute Co. Ltd., Wuhan, China; email:lila@brdi.com.cn*: corresponding author

Keywords: deep-cut canyon; wind characteristics; field measurement; non-stationary; turbulence
Abstract: The stiffness of long-span bridges across mountainous canyons is small, and wind has
a significant impact on the structure. Affected by complex terrain, wind characteristics are
significantly different between canyons and flat areas. Although there have been some studies onwind in canyons, the altitude and depth of canyons are relatively limited. There are large
temperature gradients, both horizontal and vertical, in plateau deep-cut canyons, and local thermal
drive greatly affects air flow. Existing studies are underrepresented, it is necessary to research thewind characteristics in plateau deep-cut canyons, under the comprehensive influence of complex
terrain and thermal drive. In this study, a meteorological mast was built at the bottom of the canyon,
and a long-term observation of wind was made using an ultrasonic anemometer. The drastic
variation of instantaneous wind speed causes non-stationarity to be prominent, and the stability offluctuating wind speed is studied by the run test. The applicability of the traditional stationarymodel is insufficient. Based on the discrete wavelet transform, the time-varying mean wind speed
is extracted, and the turbulence characteristics are analyzed by the non-stationary model. The
research results provide a reference for the wind resistance design of bridges in similar landforms.
1. Introduction
Accurately obtaining the wind characteristics at the bridge site is critical for the wind resistancedesign of long-span bridges. Generally, to study the wind characteristics of complex terrain bywind tunnel tests, computational fluid dynamics (CFD) and field measurements. Relatively, there
are assumptions and simplifications in tests and CFD, and measurements are more direct andefficient. Measurement of wind fileds is usually done to establish the meteorological mast or
automatic weather station at the bridge site (Li et al., 2022), and there are also researchers whoplace sensors on construction catwalks, main girders or bridge towers (Yu et al., 2019). Affectedby complex terrain, the wind speed in deep-cut canyons has strong non-stationarity, and the
traditional stationary model overestimates the turbulence characteristics of the wind (Yu et al.,2019). The top of the plateau canyon is covered with perennial snow, and the bottom is relatively
warm and dry. There are larger temperature gradients in the canyon, and local thermal drive has a69

Bridge Engineering Institute Conference 2023 (BEI-2023)
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significant impact on wind flow (Li et al., 2022). In the study, a V-shaped plateau deep-cut canyon
is selected as the research object. A meteorological mast was set at the bottom of the canyon andanemometers were installed, to obtain the mean wind characteristics, such as wind speed and winddirection. The stationarity of fluctuating wind speed was discussed by the run test, and theturbulence wind characteristics were analyzed based on the non-stationary model.
2. Canyon Topography and Measurement System
The wind field measurement is based on the construction of a long-span bridge on the QinghaiTibet Plateau. The landform of the bridge site is a typical V-shaped plateau deep-cut canyon, withan altitude of the mountaintop above 5000 m and a vertical height difference of over 2000 m. A
meteorological mast was set up at the bottom of canyon, and an ultrasonic anemometer was placedat the top of mast, with a height of 40 m and a sampling frequency of 10 Hz, as shown in Fig. 1.
N

Bridge Axis

Ultrasonic Anemometer

Bridge Site
Meteorological Mast

Model:WindMaster Pro
Height:40m
Range:0~65m/s and 0~360°
Sampling frequency:10Hz

Fig. 1. Feature of canyon topography and layout of measurement system
3. Wind Characteristics
3.1. Mean wind speed and wind attack angle
The length of the basic time interval significantly affects the results of wind characteristics, and is
uniformly taken as 10 min. Furthermore, the impact of low winds on the structure is relatively
limited, and only 1225 samples with a mean wind speed higher than 6 m/s were studied, accountingfor 2.3% of the total. The distribution of wind speed and wind attack angle is shown in Fig. 2. The
wind speed from 90° to 180° of wind direction is lower, and with the increase in wind speed, thewind direction is concentrated in the range of 315° to 0°. The wind attack angle is basicallynegative from 90° to 270° of wind direction, and positive from 270° to 90°. The attack angle issignificantly related to wind direction, and its value is more concentrated at high wind speeds.

Fig. 2. Distribution of wind speed and wind attack angle: (a) wind speed; (b) wind attack angle

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3.2. Wind speed model and stationary test
The fluctuating wind speed is obtained by subtracting its own mean value from the instantaneouswind speed. Stationary model assumes that mean wind speed remains constant, as shown in Eq. 1.

U (t=) U + u (t )(1-a)

V (t ) = v(t )(1-b)

W (t ) = w(t )(1-c)

where u(t), v(t) and w(t) denote the zero-mean fluctuating wind speeds in downwind, acrosswind
and vertical; and U is the constant mean wind speed. Non-stationary model assumes that mean
wind speed changes gradually over the basic time interval, as shown in Eq. 2.

U=(t ) U (t ) + u (t )

(2-a)

V=
(t ) V (t ) + v(t )

(2-b)

W=
(t ) W (t ) + w(t )(2-c)

where U (t ) , V (t ) and W (t ) denote the time-varying mean wind speed in downwind, acrosswind
and vertical. Here, the discrete wavelet transform is used to extract the time-varying mean windspeed, db10 wavelet is selected, and the decomposition level is 9. The run test was used to analyze
the stationarity of the fluctuating wind speed, the results as shown in Table 1. The non-stationarityof wind is stronger, decreased with the increase in wind speed. The non-stationary model cansignificantly reduce the non-stationarity of wind for the analysis of turbulence characteristics.
Table 1. Proportion of stationary fluctuating wind based on run test
0~6 m/s (Mean wind speed)
6~15 m/s (Mean wind speed)Wind speed model
u
v
w
u
v
w
Stationary
13.4%
13.0%
13.0%
30.6%
34.3%
32.6%
Non-stationary
58.6%
58.6%
58.6%
68.1
68.2%
68.8%
3.3. Turbulence intensity and gust factor
Fig. 3 shows turbulence intensity, and Fig. 4 shows gust factor. The mean values in downwind,crosswind, vertical are 0.182, 0.149, 0.164 of turbulence intensity, and 1.482, 0.411, 0.439 of gustfactor. The ratios between components are Iu:Iv:Iw=1:0.82:0.90, and Gu-1:Gv:Gw=1:0.85:0.91.
Turbulence intensity and gust factor decrease with the increase in wind speed, and become moreconcentrated. The ratios of two groups are very close, and the distribution with wind direction isbasically the same, which symbolizes the correlation between turbulence intensity and gust factor.

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Fig. 3. Distribution of turbulence intensity: (a) Iu; (b) Iv; (c) Iw

Fig. 4. Distribution of gust factor: (a) Gu; (b) Gv; (c) Gw
3.4. Turbulence integral scaleFig. 5 shows the distribution of the turbulence integral scale, and the mean values of downwind,crosswind, and vertical are 28.8 m, 22.7 m, and 20.1 m, respectively. The ratio between thecomponents is Lu:Lv:Lw=1:0.79:0.70. Turbulent eddies have larger scales at high wind speeds.

Fig. 5. Distribution of turbulence integral scale: (a) Lu; (b) Lv; (c) Lw
4. Conclusions
In this study, measurement of wind filed was carried out in a plateau deep-cut canyon. Restricted
and obstructed by terrain, wind speed is low, with only 2.3% exceeding 6m/s. The non-stationarityof wind is prominent, and the turbulence characteristics are analyzed based on the non-stationarymodel. The correlations between turbulence characteristics and wind speed or wind direction are72

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discussed, and the values and ratios between components are given. Turbulence intensity and gustfactor decrease, and the turbulence integral scale increases, with the increase in wind speed.
5. References
Li Y, Jiang F, Zhang M, Dai Y, Qin J and Zhang J. (2022) ‘Observations of periodic thermallydeveloped winds beside a bridge region in mountain terrain based on field measurement’, Journalof Wind Engineering and Industrial Aerodynamics, 225: 104996.
Yu C, Li Y, Zhang M, Zhang Y and Zhai G. (2019) ‘Observations of periodic thermally-developed
winds beside a bridge region in mountain terrain based on field measurement’, Journal of Wind
Engineering and Industrial Aerodynamics, 186, 94-104.

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Dynamical Analysis of Vehicle-Bridge Interaction System under Ice and Wind Loads
Tianyu Wu1,2,*, Wenliang Qiu1, Guowen Yao2, and Zengwei Guo2
Faculty of Infrastructure Engineering, Dalian University of Technology, Dalian City 116024, China
State Key Laboratory of Mountain Bridge and Tunnel Engineering, Chongqing Jiaotong University,
Chongqing City 400074, China
:
:

1
2

Abstract: This study constructs a complete analysis framework of ice-wind-vehicle-bridge interaction toinvestigate dynamic responses of the coupled system. Ice load, wind load, nonlinear soil spring andadditional water mass are all integrated into a full bridge model based on a sea-crossing bridge with runningvehicles. The results indicate that the combined action of ice and wind has no superimposed effect on the
movement of the bridge, but has a restraining effect. The coupled dynamic responses of the vehicle cannot
be combined by the superposition under separate ice and wind.
Keywords: Sea-crossing bridges; ice-wind-vehicle-bridge interaction; vibration analysis
1. Introduction
There is a harsh working environment for sea-crossing bridges in cold sea region. As well as being able towithstand the impact of ice load, it must also take into account the combined effect of strong wind,
water-structure interaction, soft foundation on the seabed, and other complex factors. This poses a great
threat to the safe operation of bridges. Therefore, the coupled vibration analysis of ice-wind-vehicle-bridge
interaction system is an unavoidable research topic for the construction of bridges in cold sea region.
Many research scholars have carried out targeted studies on the analysis of vehicle-bridge coupledvibration. For the ice load action, Xia[1] establishes a numerical analysis model of the vehicle-bridge coupledsystem under ice load and proposes an iterative method applicable to the time-integrated full process of thevehicle-bridge dynamic coupled system. Xia et al.[2] systematically investigates the influence law of impact
load type, impact intensity, train velocity on the safety index. Specifically, it describes the method for
determining the impact intensity-train velocity threshold curve. Wu et al.[3] defines the contact relationshipbetween the wheels and the bridge deck by using the penalty function method, and analyzes the driving
safety under the random vibration caused by ice load.
However, there is a lack of research on the vibration between vehicles and bridges under the combined
action of ice and wind. As such, this study presents a framework for the dynamic analysis of coupled
ice-wind-vehicle-bridge systems. It systematically studies the dynamic response of bridges and vehicles
under the combined action of ice and wind loads, and analyzes the effects of different ice and windvelocities on the safety of traffic on bridges. The results of this study can be used as a reference to evaluate
the driving safety of sea-crossing bridges in cold sea region.
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2. Analysis Framework of Ice-Wind-Vehicle-Bridge Interaction System
2.1. Modeling of the high-sided road truck
This study focuses on the vehicle-bridge interaction of the high-sided road truck under lateral load. A twoaxle four-wheel light high-sided road truck is adopted in this study, as shown in Fig. 1. The light truck is
composed of rigid body, spring and damper, with 13 degrees of freedom. The following generalized
displacement vector can represent the dynamical system of the light truck used in this study.
V = {Z v Yv θ v φv ψ v Z s11 Ys11 Z s12 Ys12 Z s 21 Ys 21 Z s 22 Ys 22 }(1)

where Zv is the vertical displacement of the vehicle; Yv is the lateral displacement of the vehicle; θv denotes
the pitch angle of the vehicle; φv denotes the rotation angle of the vehicle; ψv represents swing angle of thevehicle. Zs11, Zs12, Ys11 and Ys12 respectively represent the vertical displacement and lateral displacement of thetwo rigid bodies on the front axle of the vehicle. Zs21, Zs22, Ys21 and Ys22 respectively represent the verticaldisplacement and lateral displacement of the two rigid bodies on the rear axle of the vehicle body.

(a) Elevation view(b) Side view(c) Plan viewFig. 1. The high-sided road truck
2.2. Modeling of the sea-crossing bridge
The overall configuration of (5×120) m non-navigable bridge is investigated in this study. The cross section
form of the bridge girder is steel box girder section, and two-way 6 lanes are set, as shown in Fig. 2. In
bridge design, the pile-cap is located at the design water level elevation, so drift ice can affect the large
volume of the pile-cap, rather than acting on the pier or pile foundation. The overall bridge model isestablished based on ABAQUS finite element platform. The girder of the bridge is built by the spatial plateelement (S8R). The piers, pile-caps, steel pipe piles are modeled by the three dimensional beam element(B31). The dynamic action of water is simulated by the additional water mass. The soil-structure interaction(SSI) is considered by using the classical p-y nonlinear soil spring. The full bridge model of vehicle-bridge
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coupling system based on ABAQUS in this work, as shown in Fig. 3. In this study, the penalty function built
into ABAQUS is used to define contact conditions between the vehicle and the bridge, allowing theseparation and sliding between the wheel of the vehicle and the deck of the bridge.

Fig. 2. Cross section configuration of the non-navigable bridge in this study (unit: cm)

Fig. 3. Full bridge model of vehicle-bridge coupling system in ABAQUS
2.3. Coupled dynamic equation of ice-wind-vehicle-bridge interaction system
The coupled dynamic equation is established with full consideration of vehicles, bridges, additional watermass and soil-structure interaction. Ice load and wind load are applied to overall dynamic system as externalloads. Therefore, the coupled dynamic equation of ice-wind-vehicle-bridge interaction system can be
expressed as:b  Cbb +Ci Cbv   w b   K bb +K s K bv   wb   Fbb + Fice +Fbw  M bb +M w 0   w(2)
 +  + 
 =
 Sym

v   Sym
M v  w
K v   wv   Fvv +Fvw
Cv   w v   Sym

here subscripts bb, vv, and bv denote the bridge, vehicle, and vehicle-bridge interaction terms, respectively.
It should be noted that duo to there is an additional damping term Ci, the new damping term in the
motion equation includes the damping of the bridge structure and the damping caused by ice load.
3. External Excitation Loads
3.1. Ice-induced vibration model
After linearization, the multilinear relationship between ice compression strength and stress rate can beapproximately expressed as:𝜎𝜎𝑖𝑖 = 𝑏𝑏𝑖𝑖 + 𝑎𝑎𝑖𝑖 𝜎𝜎𝚤𝚤̇

where ai and bi are slope and intercept of segment i. σi is the crushing strength of ice.
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Määttänen[4] gives the expression of stress rate and relative velocity during the compressive failure ofice.

where D is the structure diameter.

(4)𝜎𝜎𝚤𝚤̇ = 8𝜎𝜎1 𝑣𝑣𝑟𝑟 /𝜋𝜋𝜋𝜋

Therefore, the total ice force on the structure can be written as:
8𝜎𝜎

8𝜎𝜎

𝐹𝐹(𝑣𝑣𝑟𝑟 ) = 𝐹𝐹(𝑣𝑣𝑖𝑖𝑖𝑖𝑖𝑖 − 𝑢𝑢̇ ) = 𝐴𝐴 × 𝜎𝜎(𝑣𝑣𝑖𝑖𝑖𝑖𝑖𝑖 − 𝑢𝑢̇ ) = 𝐴𝐴 �𝑎𝑎𝑖𝑖 × (𝑣𝑣𝑖𝑖𝑖𝑖𝑖𝑖 − 𝑢𝑢̇ ) 𝜋𝜋𝜋𝜋1 + 𝑏𝑏𝑖𝑖 � = �𝐴𝐴 × 𝑏𝑏𝑖𝑖 + 𝑣𝑣𝑖𝑖𝑖𝑖𝑖𝑖 × 𝐴𝐴 × 𝑎𝑎𝑖𝑖 𝜋𝜋𝜋𝜋1 � −
8𝜎𝜎

(5)

𝑢𝑢̇ × 𝐴𝐴 × 𝑎𝑎𝑖𝑖 𝜋𝜋𝜋𝜋1

where h is the ice thickness, A is the contact area under ice-induced vibration.
For the linearized ice load model, the ice load simulation can be divided into three piecewise functions
according to the ice stress rate, which can be written as: b + υ × a 8σ 1
A
 1 ice 1Dπ b + υ × a 8σ 1
 2 ice 2
π D  A
F(vice) =

(6)

b3 A
The damping coefficient changes with the stress rate of ice, and its subsection expression can be written as:
a1
C’ =

a2

8σ 1
8σ 1

πDA
(7)

πDA

0

3.2. Wind load
3.2.1. Wind load on bridges
The steady wind load on the girder of the bridge includes the lift force (Lst) in the vertical direction, the dragforce (Dst) in the lateral direction and the bending moment (Mst) in the torsion direction, which can be
expressed as:
1

(8a)

1

(8b)

1

(8c)

𝐿𝐿 = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵𝐶𝐶𝐿𝐿

𝐷𝐷 = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵𝐶𝐶𝐷𝐷

𝑀𝑀 = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵 2 𝐶𝐶𝑀𝑀

here U denotes the average wind velocity; B stands for the width of the girder; CD, CL and CM represent thedrag, lift and moment coefficients.
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The buffeting force including aerodynamic admittance function of the bridge girder suggested by
Scanlan[5] can be expressed as:
1

2𝑢𝑢(𝑡𝑡)1

2𝑢𝑢(𝑡𝑡)

𝐿𝐿𝑏𝑏 (𝑡𝑡) = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵 �𝐶𝐶𝐿𝐿 𝐼𝐼𝐿𝐿𝐿𝐿

𝐷𝐷𝑏𝑏 (𝑡𝑡) = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵 �𝐶𝐶𝐷𝐷 𝐼𝐼𝐷𝐷𝐷𝐷
1

𝑈𝑈

𝑀𝑀𝑏𝑏 (𝑡𝑡) = 2 𝜌𝜌𝑈𝑈 2 𝐵𝐵 2 �𝐶𝐶𝑀𝑀 𝐼𝐼𝑀𝑀𝑀𝑀

𝑈𝑈+ [𝐶𝐶𝐿𝐿′ + 𝐶𝐶𝐷𝐷 ]𝐼𝐼𝐿𝐿𝐿𝐿+ 𝐶𝐶𝐷𝐷′ 𝐼𝐼𝐷𝐷𝐷𝐷

2𝑢𝑢(𝑡𝑡)𝑈𝑈

𝑤𝑤(𝑡𝑡)𝑈𝑈′+ 𝐶𝐶𝑀𝑀
𝐼𝐼𝑀𝑀𝑀𝑀�

𝑤𝑤(𝑡𝑡)

𝑤𝑤(𝑡𝑡)
𝑈𝑈

𝑈𝑈

�

�

(9a)
(9b)
(9c)

in which Lb(t), Db(t) and Mb(t) are the buffeting lift, drag and moment; CL′, CD′ and CM′ are the slope of the
CL, CD and CM; u(t) and w(t) denote the velocity components of fluctuating wind in horizontal and verticaldirection; I( ) stands for the aerodynamic admittance function.
3.2.2. Wind load on vehicles
The aerodynamic wind loads on the vehicle can be calculated based on the quasi-steady approach whose
formulae is determined by a function of the yaw angle ψ[6]. The aerodynamic load and moment on thevehicle can be calculated according to the following equations:
1

(10a)

1

(10b)

1

(10c)

𝐹𝐹𝑥𝑥 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝐷𝐷 (Ψ)𝐴𝐴𝑓𝑓
𝐹𝐹𝑦𝑦 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝑆𝑆 (Ψ)𝐴𝐴𝑓𝑓
𝐹𝐹𝑧𝑧 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝐿𝐿 (Ψ)𝐴𝐴𝑓𝑓
1

(10d)

1

(10e)

1

(10f)

𝑀𝑀𝑥𝑥 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝑅𝑅 (Ψ)𝐴𝐴𝑓𝑓 ℎ𝑣𝑣

𝑀𝑀𝑦𝑦 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝑃𝑃 (Ψ)𝐴𝐴𝑓𝑓 ℎ𝑣𝑣
𝑀𝑀𝑧𝑧 = 2 𝜌𝜌𝑈𝑈𝑅𝑅2 𝐶𝐶𝑌𝑌 (Ψ)𝐴𝐴𝑓𝑓 ℎ𝑣𝑣

here Fx stands for drag force, Fy stands for side force, Fz stands for lift force, Mx stands for rolling moment,
My stands for pitching moment and Mz stands for yawing moment; CD(ψ) denotes the drag force coefficient,CS(ψ) denotes the side force coefficient, CL(ψ) denotes the lift force coefficient, CR(ψ) denotes the rolling
moment coefficient, CP(ψ) denotes the pitching moment coefficient, CY(ψ) denotes the yawing moment
coefficient; Af denotes the reference area; hv denotes the reference height.
4. Dynamic Responses Analysis
4.2. Influence of combined ice and wind loads
It can be seen from Fig. 4 that the maximum lateral displacement of the bridge girder under ice load isgreater than that under combined ice and wind loads. Although the ice and wind loads are imposed along thesame direction of the transverse bridge, the lateral dynamic response of the bridge is not a simple linear
superposition of the displacement caused by ice and wind loads. It can be interpreted as when the ice load isapplied to the pile-cap of the bridge substructure, the movement direction of the girder of the bridge
superstructure is opposite to the application direction of the ice load. The wind load is directly applied to the78Bridge Engineering Institute Conference 2023 (BEI-2023)
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bridge girder, and the motion direction of the bridge girder under wind load is the same as the application
direction of the wind load. Therefore, the combined action of ice and wind loads have no superposition
effect on the movement of the bridge girder, but has a restraining effect in the opposite direction. In addition,
it can be found that the maximum displacement of the bridge girder decreases with the increase of windvelocity. For example, when the wind velocities are 0 m/s, 10 m/s, 15 m/s and 20 m/s, the maximum lateraldisplacement are 62.75 mm, 59.46 mm, 57.51 mm and 55.05 mm, respectively. This again shows that thewind load and ice load have opposite contributions to the motion of the bridge girder. However, from theresults, ice load has a greater impact on the dynamic response of the bridge girder than wind load.
It can be seen from Fig. 5 that the vertical displacement of the bridge girder under ice load is greaterthan that of under combined ice and wind loads. It can be interpreted as the ice load is only applied to thetransverse bridge, which has little contribution to the vertical vibration of the bridge. However, wind load on
the bridge includes the lifting force applied on the bridge girder, which can reduce the vertical displacementof the bridge girder. It can also be found that due to the increase of wind velocity, the lifting force on thebridge girder also increases, which makes the vertical displacement of the bridge girder show a decreasing
trend.
It can be clearly seen from Fig. 6 that when only ice load acts on the bridge, the lateral contact force ofthe vehicle fluctuates at a lower average level. This is because when there is no transverse wind load acting
on the bridge, the vehicle movement only occurs with the transverse vibration of the ice-induced bridge.
However, when considering the combined effect of ice and wind loads on the bridge, it can be found that thelateral contact force of the vehicle increases significantly and fluctuates at a higher average level. For
example, when the wind velocities are 0 m/s, 10 m/s, 15 m/s and 20 m/s, the average value of lateral contact
force are 1.75 kN, 9.09 kN, 12.91 kN and 17.77 kN, respectively. This also shows that with the increase ofwind velocity, the lateral contact force of the vehicle will also increase, which is very unfavorable to thedriving safety.
It can be seen from Fig. 7 that the vertical contact force of the vehicle under the combined ice and windloads is slightly larger than that under ice load. The vertical contact force of the vehicle is mainly affected by
the pavement roughness, but the wind load also has a certain impact on the vertical contact force of thevehicle. With the increase of wind velocity, the vertical contact force of the vehicle increases slightly.
Lateral displacement [m]

0.10

Ice
Ice+Wind (vw1=10m/s)
Ice+Wind (vw2=15m/s)
Ice+Wind (vw3=20m/s)

0.05
0.00
-0.05
-0.10
0.0

0.1

0.2

0.3

0.4

0.5

x/L

0.6

0.7

0.8

0.9

1.0

Fig. 4. Lateral displacement of bridge girder under combined ice and wind loads
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Vertical displacement [m]

0.004

Ice
Ice+Wind (vw1=10m/s)
Ice+Wind (vw2=15m/s)
Ice+Wind (vw3=20m/s)

0.002
0.000
-0.002-0.004-0.006
0.0

0.1

0.2

0.3

0.4

0.5

x/L

0.6

0.7

0.8

0.9

1.0

Fig. 5. Vertical displacement of bridge girder under combined ice and wind loads
Lateral contact force [kN]30

Ice
Ice+Wind (vw1=10m/s)
Ice+Wind (vw2=15m/s)
Ice+Wind (vw3=20m/s)

20
10
0
-10
-20
-30
-40
0.0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1.0

x/L

Fig. 6. Lateral contact force between tires and bridge deck under combined ice and wind loads
Vertical contact force [kN]

-20

Ice
Ice+Wind (vw1=10m/s)
Ice+Wind (vw2=15m/s)
Ice+Wind (vw3=20m/s)

-30
-40
-50
-60
-70
0.0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1.0

x/L

Fig. 7. Vertical contact force between tires and bridge deck under combined ice and wind loads
5. Conclusions
The ice load has a greater influence on the lateral dynamic response of the bridge, while the wind load has amore significant influence on the lateral dynamic response of the vehicle. Due to the different action
characteristics of ice load and wind load on the vehicle-bridge interaction system, the lateral dynamic
responses of the bridge and vehicle under separate ice and wind loads cannot be combined by superposition.
The combined action of ice and wind should be reasonably considered in the driving safety analysis ofsea-crossing bridges in ice covered waters.
6. Acknowledgments
This work was financially supported by the National Natural Science Foundation of China (52201313), theFundamental Research Funds for the Central Universities (DUT22RC(3)091), State Key Laboratory Science
Foundation of Mountain Bridge and Tunnel Engineering (SKLBT-2101).
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7. References[1] Xia C Y. Dynamic responses of train-bridge system subjected to collision loads and running safety
evaluation of high-velocity trains. Doctoral Dissertation, Beijing Jiaotong University, 2012.
[2] Xia C Y, Lei J Q, Zhang N, et al. Dynamic analysis of a coupled high-velocity train and bridge system
subjected to collision load. Journal of Sound and Vibration, 2011, 331(10): 2334-2347.
[3] Wu T Y, Qiu W L, Kim C W, Chang K C, Lu X Z. Dynamic responses of a vehicle-bridge-soil
interaction system subjected to stochastic-type ice loads. Structure and Infrastructure Engineering, 2021,
2023586: 1-20.
[4] Määttänen M. On conditions for the rise of self-excited ice-induced autonomous oscillations in slendermarine pile structures. Winter Navigation Research Board, 1978.
[5] Scanlan R H, Jones N P. Aeroelastic analysis of cable-stayed bridges. Journal of Structural Engineering,
1990, 116(2): 279-297.
[6] Zhou Y and Chen S. Fully coupled driving safety analysis of moving traffic on long-span bridges
subjected to crosswind. Journal of Wind Engineering and Industrial Aerodynamics, 2015, 143:1-18.

81

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Safety Analysis of Wind Barriers on the Bridge under the 400 km/h
High-Speed Trains Loading of Slipstreams
Dianyi Guo1, Yunfeng Zou2*, Xuhui He3, and Haobo Liang41: School of Civil Engineering, Central South University, Changsha and China; email:guodianyi@csu.edu.cn
2: School of Civil Engineering, Central South University, Changsha and China; email:yunfengzou@csu.edu.cn
3: School of Civil Engineering, Central South University, Changsha and China; email:xuhuihe@csu.edu.cn
4: School of Civil Engineering, Central South University, Changsha and China; email:haoboliang0816@163.com*: corresponding author

Keywords: high-speed railway; loading of slipstreams; wind barrier; dynamic response;
concrete damage
Abstract: The loading of slipstreams induced by a high-speed train (HST) passing through thebridge at 400 km/h may generate repeated impacts on the wind barrier, which can causesignificant dynamic responses and structural damage, such as concrete cracking at the foundation.
This poses a threat to the long-term durability of the barrier itself. To accurately evaluate theeffect of slipstreams on the wind barrier, this paper conducts a three-step study on the dynamic
response of the wind barrier column structure and the structural damage of the anchor-end
concrete under the loading of slipstreams. Firstly, numerical simulation is used to analyze thevariation of slipstream loading on the wind barrier when a 400 km/h HST passes over the bridge,
and the distribution and laws of wind pressure inside the barrier for trains are analyzed,
determine the loading of slipstreams curve at the most disadvantageous position. Next, finite
element analysis(FEA) is used to study the structural static and dynamic response of the windbarrier column under the slipstream loading, considering the influence of different wind barrierheights, column spacing, and distance between the center-line of the track and the wind barrier.
Finally, nonlinear finite element analysis (NFEA) is used to calculate the degree of damage tothe anchor-end concrete, the safety and durability of the wind barrier on the bridge when a 400km/h HST passes over it are evaluated.
1. Introduction
With the further increase in the speed of high-speed railways, the 400 km/h high-speed railway
will be put into use in the future. Compared with ground-level travel, bridge-level travel has thecharacteristics of high wind speed and prominent aerodynamic interference of the vehicle-bridge,
and the operation of trains in high wind conditions will be limited or suspended. Previous studies
have shown that the wind speed threshold for a train passing over a 32m simply supported beam
bridge at a speed of 400 km/h is 12.5m/s (Li et al. 2021). Wind barriers installed on both sides of82

Bridge Engineering Institute Conference 2023 (BEI-2023)
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the bridge are currently the most common protection measure, which can create a lower local
wind environment to protect the operation of trains. However, high-speed trains will cause
fluctuating pressure loads on the surrounding structures, resulting in structural deformation, boltloosening, and other disasters (Carassale and Brunenghi, 2013). Currently, many scholars havefocused on the vibration response of wind barriers. Belloli et al. (2012) conducted on-siteexperiments and wind tunnel tests to obtain the pressure load curve of train wind loads on soundbarriers. They established a finite element model of sound barriers and calculated their vibration
response. Tokunaga et al. (2013; 2016) conducted on-site measurements and numericalcalculations on half-covered snow-type sound barriers. The results showed that the dynamic
response of the sound barrier may resonate with the vibration frequency of the train, which is thesuperposition of resonance and dynamic effects of the train wind. Carassale and Brunenghi(2013) conducted on-site measurements of steel frame structures beside the track, measured theacceleration response of the structure, and established a mathematical model to predict the
dynamic behavior of the structures. Rocchi et al. (2018) conducted on-site measurements tostudy the dynamic loads of different trains on sound barriers and measured the strain of thesound barrier base. Friedl et al. (2011) established a finite element model of sound barriers andanalyzed the response of the structure under dynamic cyclic loads, obtaining the maximum
deformation and stress on a single column, and finally calculated the fatigue performance of thesound barrier. Vittozzi et al. (2017) modeled upright barriers and conducted dynamic analysis
with tuned mass dampers (TMD) installed at the top of the column. The results showed that the
TMD system can effectively reduce vibration amplitude and cycle times, and reduce the stressand fatigue of the sound barrier base. Many scholars mainly focus on the dynamic response ofwind barriers themselves, and rarely pay attention to other safety issues of wind barriers, such asconcrete damage. This will greatly affect the estimated service life of wind barriers and pose athreat to the safety of train operation. Therefore, research on the safety of wind barriers for 400km/h high-speed railway bridges is urgently needed.
2. Methodology
2.1 Computational fluid dynamics
This study adopts the “head car + middle car + tail car” formation mode and uses a combinationof structured and unstructured grids to accurately simulate the detailed structures of the train andthe barrier. The accuracy of the simulation is verified by comparing the results with the fieldmeasurements provided by (Xiong et al. 2018). The ultimate goal is to calculate the fluctuating
wind pressure on the wind barrier under the slipstreams generated by a 400 km/h high-speed
train.
2.2 FEA dynamic response loading
The dynamic response analysis of the wind barrier to the slipstream loading is carried out using
the full method, which takes into account the train’s travelling wave effect. The loading ofslipstreams induced by HST are applied to the wind barrier, as shown in Fig. 1, which illustrates
the lateral force distribution pattern on the wind barrier when train reaches same locations.
2.3 NFEA concrete damage calculation
We established a finite element model of the wind barrier anchor and concrete using ABAQUS,
and calculated the stress and deformation of each component using 8-node reduced integration
solid elements (C3D8R). The concrete used the Mohr-Coulomb elastic-plastic constitutive model,

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while other materials used linear elastic models. The structure of the finite element model isshown in Fig. 2. We applied dynamic response analysis to calculate the maximum displacement
value of the wind barrier column and then determined the concrete damage.

The loading ofslipstreams

Fig. 1. Lateral force distribution pattern on the wind barrier when train reaches same locations(train moving left): (a)the train didn’t arrive; (b)headstock arrive; (c)tailstock arrive
50
2.7

60

40

25

Fig. 2. Model grid (Unit: cm)
3. Results and Discussion
3.1 Slipstream effects of the wind barrier
From Fig. 3, it can be observed that as the train passes through the wind barrier area, the windpressure on the barrier gradually decreases from bottom to top along the height direction,
showing a trend of gradually decreasing. Along the length direction, it shows a pattern of larger
pressure in the middle and smaller pressure at the entrance and exit.
(b) 800(a) 600600

200

Fluctuating pressure/Pa

Fluctuating pressure/Pa

400

Positive pressure peak
Negative pressure peak

0

-200

-400

-600
0.0

400

Positive pressure peak
Negative pressure peak

200
0
-200
-400
-600

0.5

1.0

1.5

2.0

2.5

3.0

-800

-2

0

2

4

6

8

10

12

14

16

18

Length from entrance/m

Height/m

Fig. 3. Wind barrier pressure peak variation curve: (a) along the height direction; (b) along thelength direction84

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3.2 Dynamic response of the wind barrier
Fig. 4 shows the dynamic response of the wind barrier, and Fig. 4 (a) shows the displacement
response of the wind barrier. It can be seen that the peak displacement response of the windbarrier is basically consistent with the “head wave” and “tail wave” of the train pulsating windload. The displacement response generated by the head wave is larger, with a maximum responseof 1.7mm. Fig. 4 (b) shows the stress response of the wind barrier, which is also basicallyconsistent with the slipstream loading. The maximum equivalent stress occurs at the bottom of
the column, resulting in stress concentration.
(a) 2.0

1.0
0.5
0.0
-0.5
-1.0

24

Von Mises stress response time history

20
18
16
14
12
10
8
6
4

-1.5
-2.0
0.0

26
22

Von Mises stress/MPa

Displacement/mm(b)

Displacement response time history

1.5

2
0.5

1.0

1.5

2.0

2.5

3.0

0
0.0

0.5

1.0

1.5

2.0

2.5

3.0

time/stime/sFig. 4. Wind barrier response time history: (a) displacement; (b) Von Mises stress
3.3 Concrete damage of the anchor-end
Fig. 5 shows the stress state of the concrete at the anchor-end. Fig. 5 (a) shows the maximum
principal tensile stress, with a maximum value of 1.97. and Fig. 5 (b) shows the tensile damage
factor of the concrete, with a maximum damage factor of 0.969.
(b)

(a)

Fig. 5. Concrete stress state: (a) maximum principal stress; (b) tensile damage
4. Conclusions
According to the slipstreams loading that the wind barrier is subjected to when the high-speed
train passes through the bridge, as well as the dynamic response of the wind barrier and thedamage to the concrete at the anchor-end, it can be concluded that the slipstream loading fromthe passing train has a significant impact on the wind barrier, and the maximum displacement
response of the wind barrier occurs at the top of the second pillar at the entrance. Additionally,
there is a risk of damage and cracking to the concrete at the anchorage end of the wind barrier, so
it is necessary to pay attention to the safety of the concrete at the anchor-end.

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5. References
Xiong X.H., Li A.H., Liang X.F., and Zhang J., 2018. Field study on high-speed train induced
fluctuating pressure on a bridge noise barrier. Journal of Wind Engineering and Industrial
Aerodynamics. 177, 157-166.
Soper D., Gillmeier S., Baker C., Morgan T., Vojnovic L., 2019. Aerodynamic forces on railway
acoustic barriers. Journal of Wind Engineering and Industrial Aerodynamics. 191, 266-278.

86Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Fatigue Analysis of Anchored Concrete Structures of Catenary Induced by
Train Wind Load Considering Different Bolt Pre-tightening
Zhuole Shu1*, Pengru Deng1,2, Hanfeng Wang1,2, and Xuhui He1,2
1: Department of Civil Engineering, Central South University, Changsha, Hunan, China; email:shuzhoule@csu.edu.cn
2: National Engineering Research Center of High Speed Railway Construction Technology,
Central South University, Changsha, Hunan, China; email: pengrudeng@csu.edu.cn*: corresponding author

Keywords: tunnel; catenary; anchored concrete structures; finite element analysis; fatigue
Abstract: The ancillary facilities in the railway tunnel are mainly affected by the wind load of
the high-speed train which may be the cause of the cracks and deteriorations commonly observedin the concrete structure at the anchorage end. In this paper, nonlinear finite element analysis(FEA) is conducted on the concrete anchorage end of the catenary with different levels of bolt
pre-tightening forces to calculate the structural stress fluctuation ranges when a train passing
through a tunnel with high-speed beforehand, where the wind load is obtained fromcomputational fluid dynamics (CFD) and the corresponding dynamic effect is included.
Hereafter, based on the quasi-static explicit analysis method which has a small computational
cost in the case of short time increments, the concrete cumulative damage caused by cyclic highspeed train wind loads and the fatigue life is calculated using the stress ranges of anchoredconcrete with different pre-tightening forces. The analysis results show that the tensile damage ofconcrete at the bolt hole interface mainly propagates in the depth direction with the increase of
the number of cycles, and the minimum fatigue life of concrete under cyclic wind load is127,500 under three bolt pre-tightening levels.
1. Introduction
Owing to the limitation of test cost and time, the mechanical research on the anchored ancillaryfacilities in tunnels usually focuses on the strength of steel members under static loads (Li et al,2017). However, the structural damage is normally initiated from the anchorage end system
which is mainly composed of lining concrete and anchor rods in the form of local adhesion
failure between rods and concrete or the development of concrete micro-cracks. As a result, theexisting analysis tends to be relatively conservative. Moreover, the catenary is a cantileverstructural with a relatively large windward area in the anchored accessary structures of a tunnel,
the anchorage structure should be more prone to the cracking and damage under the action of
cyclic wind load.
Therefore, fatigue analysis of the catenary in the tunnel under the high-speed train windload was carried out focusing on the cracking and damage in the anchorage structures in thisstudy. Firstly, CFD and structural dynamic calculation are conducted to calculate the train windload imposed on the catenary and the amplification factor related to the dynamic effect. On this87Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

basis, the nonlinear FEA method is used to calculate the damage of the anchorage concrete underthe bolt pre-tightening, self-weight and working load, and train wind load. Furthermore, a new
way of calculating the fatigue life of concrete by using the quasi-static explicit analysis is
proposed.
2. MethodAt first, a CFD model include train, tunnel and catenary with a ratio of 1:1 was established in
FLUENT, it is shown in Fig. 1(a) and used to calculate the critical train wind load on thecatenary when the train passed through a tunnel with a length of 800m at a speed of 350km/h.
Secondly, the finite element static and dynamic analysis of the catenary structure containing thecantilever bracket (Qi et al, 2018) was carried out to calculate the dynamic amplification factor
under the wind load. In the FEA calculation, the static implicit analysis based on ABAQUS wasused to calculate the structural stress and damage of the model shown in Fig. 1(b) under bolt pretightening force, self-weight load and the single wind load. The nonlinear behavior of concrete in
tension and compression was represented by the concrete damage plasticity (CDP) model. In
order to simplify the boundary conditions in engineering, the contact between the steel plate of
the catenary davit and the concrete was defined as “friction”, and the contact between the anchor
bolts, and concrete, base plate was defined as “bonded”. The yield strength and elastic modulusof the steel plate and bolts are defined as “fy=235 MPa, Es=210GPa”, and those of concrete were
defined as “fc=20.1MPa, ft=2.01 MPa, Es=30 GPa”. In addition, the fatigue analysis of concrete
was based on the quasi-static explicit calculation method, using the static calculation results asinitial conditions to calculate the cumulative damage of the same model under cyclic wind loads,
and the maximum number of given wind load cycles is 2 million.
(b)

Y

NP
Z

Z

8500

NR
FR

Y
X

BondedFP

Y

Cross section

O1

Z

Tangentia
friction

300

2050

(a)

2170

2590

O2

BondedFixed boundary

9000

Note: NP means the bolt that near the train passing side(X-), and positive side of cantilever bracket(Y+)
NR means the bolt that near the train passing side(X-), and reverse side of cantilever bracket(Y-)
FP means the bolt that far the train passing side(X+), and positive side of cantilever bracket(Y+)
FR means the bolt that far the train passing side(X+), and reverse side of cantilever bracket(Y-)
Bolt diameter: 20mm, Anchoring depth: 170mm

Fig. 1. Numerical models (mm): (a) dimensions of train-tunnel-catenary cross section; (b)boundary conditions of the anchorage structure88Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

3. Results and Discussions
3.1 Structural behaviors of the concrete anchorage structure under once train wind load
Using the time history of the wind load calculated by CFD the structural dynamic amplification
factor is calculated to be about 1.358. Correspondingly, the equivalent force and bending
moment applied on the FE model of anchorage end structure at the defined origin of coordinates
is calculated as shown in Table 1
Table 1. Loads applied to the FE modelType
Direction and Value
Bolt pre-tightening
FZ = 27272 N
The critical train wind load
FX =664.28 N，M Y =⋅855.58 m，M Z =−392.33 N ⋅ mLoad step
1
2

The bolt pre-tightening force in this model is calculated as follows:
FT =

T
KD

(1)

where T is the tightening torque, defined as 120, 140 and 160N·m. K is the dimensionlesscoefficient of the tightening torque, defined as 0.22. D is the diameter of the bolt, defined as
0.02m (Kong et al, 2022).
From FEA, the largest equivalent stress of steel structure appears at the NP bolt, which is137.68MPa, and it is less than the fatigue limit (181.1MPa) of this type of steel, steel structures
are considered to have a fatigue life greater than 10 million under this cyclic wind loads. In
addition, the distribution of concrete damage factors along the depth near each bolt hole iscalculated using the tensile and compressive stress distributions as shown in Fig. 2.
1.0
NP
FP
NR
FR

0.8
0.624

0.6

Compression damage factor

Tensile damage factor

1.0

0.4

0.2

0.0

43.67
0

20

40

0.6
0.4

0.325

0.2
11

76.98
60

NP
FP
NR
FR

0.8

80

100

120

140 160

180

Depth along the hole (mm)

0.0

0

20

40

60

80

100 120 140 160 180

Depth along the hole (mm)

(a)

(b)

Fig. 2. Damage factor distributed along depth: (a) tensile damage; (b) compression damage
It can be found that the interface concrete at the bolt hole NP has a peak value of tensile damage
factor of approximate 0.8. Referring the critical damage factors for judging the tensile and89

Bridge Engineering Institute Conference 2023 (BEI-2023)
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compressive failure of concrete, i.e., 0.624 and 0.325, as proposed by (Zhou 2019), the depth of
concrete failure is calculated to be about 77mm.
3.2 Fatigue behaviors of the concrete anchorage structure under cyclic train wind loads
The tensile damage distribution of the concrete near the NP hole under pre-tightening torque of
120 N·m and the cyclic train wind load is shown in Fig. 3(a). It was found that the fatigue failureof the concrete at the bolt interface is mainly caused by the propagation of the tensile damage
factor in the depth direction, and the damage peak value is 0.83. For this reason, the fatigue livesof concrete with NP holes under three tightening torque were 312,500, 205,000, and 127,500. As
shown in Fig. 3(b), the tensile damage of concrete rapidly deepens in the middle of the bolt as
the bolt torque increases, and the damage distribution under the three torques is approximately
the same when the anchorage concrete fatigue fails.
1.0

Tensile damage factor(a)Location: NP
Pre-tightening Torque: 120N·m
Cycle=0

100,000

312,500

(b)

0.8
0.624

0.6
120-0
120-312,500
140-0
140-205,000
160-0
160-127,500

0.4

0.2

0.0

0

20

40

60

80

100

120

140 160

180

Depth along the hole (mm)Fig. 3. Distribution of tensile damage factors: (a) different cycles; (b) different depths
4. Conclusions
In this paper, a quasi-static explicit analysis method is proposed to calculate the damage
accumulation of anchored concrete, and it is found that the results of the concrete fatigue failure
at the bolt hole interface is due to the rapid deepening of the tensile damage. Affected bydifferent bolt pre-tightening levels, the fatigue life of concrete under the wind load is calculatedwith a minimum value of 127,500 cycles and a maximum value of 312,500 cycles.
5. References
Li, R., Zhang, W., Ning, Z., Liu, B., Zou, D., & Liu, W. (2017). Influence of a high-speed train
passing through a tunnel on pantograph aerodynamics and pantograph–catenary interaction.
Proceedings of the Institution of Mechanical Engineers, Part F: Journal of Rail and Rapid
Transit, 231(2), 198-210.
Qi, G. F., Xiao, X. H., Zhao, H., & Chen, J. F. (2018). Fatigue life evaluation on key components
of high-speed railway catenary system. Journal of Central South University, 25(8), 1958-1965.

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Xiao, Y., Chen, Z., Zhou, J., Leng, Y., & Xia, R. (2017). Concrete plastic-damage factor for
finite element analysis: Concept, simulation, and experiment. Advances in Mechanicalengineering, 9(9), 1687814017719642.
Zhou, Y., Shi, S., Cai, Q., Zhang, Y., Liang, J., & Chen, Y., (2019). Damage model of anti-slide
pile based on energy loss and its application. Journal of Geomechanics, 25(6), 1107-1115. (inChinese)Kong, Q., Li, Y., Wang, S., Yuan, C., & Sang, X. (2022). The influence of high-strength bolt
preload loss on structural mechanical properties. Engineering Structures, 271, 114955.

91

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Influence of Spacing on Aerodynamic Interference of Parallel Rail-Cum-Road
Bridges with Twin Separated Parallel Girders
Jing He1, Yunfeng Zou2*, Xuhui He3, and Lulu Liu4
1

: School of Civil Engineering, Central South University, Changsha and China; email:214811083@csu.edu.cn
2: School of Civil Engineering, Central South University, Changsha and China; email:yunfengzou@csu.edu.cn
3: School of Civil Engineering, Central South University, Changsha and China; email:xuhuihe@csu.edu.cn
4: School of Civil Engineering, Central South University, Changsha and China; email:18737198956@163.com*: corresponding author

Keywords: parallel rail-cum-road bridges with twin separated parallel girders; numericalsimulation; wind tunnel test; aerodynamic disturbance; flow field mechanism
Abstract: Main girder spacing is one of the main factors affecting the aerodynamic interference
effect of double parallel bridges. In order to study the effect of main girder spacing on the
aerodynamic disturbance effect of the parallel asymmetric double bridges, based on 2D unsteadyReynolds averaged Navier-Stokes (URANS) simulations, and the wind pressure distribution andflow field of asymmetric double-width girders in the gap-width ratios D/B (0.25 to 10) areverified by wind tunnel tests based on the numerical calculation results. The comparative
analysis of wind pressure distribution and flow field characteristics of asymmetric double-width
girders in the range of gap-width ratios D/B was carried out. The results show that when thespacing D/B <2.0, the influence of aerodynamic disturbance on both upstream and downstream
girders is more obvious, and the downstream girder becomes a single flow state because it iscompletely wrapped in the shear layer formed by the upstream highway main girder, which leads
to a more significant effect of the aerodynamic disturbance. When the spacing is D/B = 5.0, theupstream wake is gradually dissipated in the gap, and the flow pattern of both graduallyconverges to that of a standalone girder, thus the aerodynamic interference gradually decreases.
1. Introduction
Large-span bridges arranged in a parallel configuration are favored by the engineeringcommunity because of their good capacity. The aerodynamic characteristics of double-width
bridges are more complicated due to aerodynamic interference. The interference of various
influencing factors is difficult to be accurately measured and described, so it is necessary to
study the aerodynamic interference effect of double-width bridges in various ways. Many
scholars have studied the influence of spacing on the aerodynamic characteristics of doublewidth bridges through wind tunnel tests or numerical simulations, and found that the downstream
main girder is more significantly affected, and summarized the law of the change of the force92

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

coefficient with spacing and wind angle of attack (Kwok, 2012; Laima, 2015). Chen et al (2014)
used a high-resolution particle imaging velocimetry (PIV) system to measure and quantify theevolution of the non-constant vortex structure around double box decks and showed that the sizeof the vortex is affected by the spacing ratio. He et al (2021) used three-dimensional LES
simulations to investigate the aerodynamic interference of parallel box girders with different
spacing, and the results showed that the critical spacing for the interference is G/B=0.25 (where
the flow has a large variation) and G/B=5.0 (greater than this spacing, the interference tends to 0).
In summary, most of the relevant research objects are symmetrical double-width bridges with thesame cross-sectional form, but less research has been done for parallel double-width bridges withdifferent main girder cross-sections. Therefore, research on the s parallel rail-cum-road bridges
with twin separated parallel girders is worthwhile.
2. Methodology
2.1.
Numerical simulation settings
The cross-sectional form is appropriately simplified, and the influence of the bridge
appurtenances such as railings and maintenance vehicle tracks are not considered. The model
scales as 1:100, and the calculation domain is set as a rectangle of 10H×25B (H=0.45m,
B=0.38m), and D is the net spacing between the two bridges, as shown in Fig. 1.

Fig. 1. Schematic representation of the computational domain
2.2.
Numerical calculation validation
Three sets of grids (40W, 60W, 80W) were selected for grid irrelevance and checked for
accuracy against the experimental results. 60W grid result fits well with the mean wind pressure
coefficients of the wind tunnel tests, and the numerical simulation method is feasible.
3. Results and Discussion
3.1.
Average wind pressure coefficient
As shown in Fig. 2, where the superscript “P” represents the position of the girders, thesuperscript “S” represents the standalone girder, and the superscript “D” represents the doublewidth girders. When D/B <2, the average wind pressure coefficient decreases with increasing
spacing, but is always larger than that of the standalone girder; while D/B>2, the impact ofaerodynamic interference upstream highway girder is smaller, and the wind pressure of the
upstream highway girder is very close to that of the standalone girder.

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0.4

0.5

P17

P34

-0.5 -0.4 -0.3 -0.2 -0.1 0.0 0.1 0.2 0.3 0.4 0.5

1.00
0.75
0.50
0.25
0.00
-0.25
-0.50
-0.75
-1.00

x/B

0.2

Cp

(b) 1.25

G/B=0.25
G/B=0.5 0.1
G/B=1.0
G/B=1.5 0.0
G/B=2.0
G/B=5.0 -0.1
0.3
G/B=10.0
standalone

Cp

1.00
0.75
0.50
0.25
0.00
-0.25
-0.50
-0.75
-1.00

0.0

0.0

-0.2

-0.2

-0.4

17

34

-0.6
P17 at D
P17 at S

-0.8
-1.0

0

1

2

3

4

5

0.0
-0.1
-0.2
0.3

0.4

G/B

6

7

8

9

x/B

P84 at D
P84 at S

P59 at D
P59 at S

-0.4

59

-0.8
-1.0

10

P84

-0.5 -0.4 -0.3 -0.2 -0.1 0.0 0.1 0.2 0.3 0.4 0.5-0.6

P34 at D
P34 at S

0.5

P59

0.2

Cp

Cp

(a) 1.25

0

1

2

3

4

84

5

G/B

6

7

8

9

10

Fig. 2. Average air pressure coefficient on highway girder: (a) upper surface; (b) lower surface
In Fig. 3, the absolute value of wind pressure at the measurement points increases withincreasing spacing. The wind pressure of the downstream railway girder is still different fromthat of the standalone girder when the spacing become larger, which shows that the aerodynamic
interference of the spacing is more significant for the downstream girder.
(a) 1.25

-0.3

P1

P17

-0.5 -0.4 -0.3 -0.2 -0.1 0.0 0.1 0.2 0.3 0.4 0.5

1.00
0.75
0.50
0.25
0.00
-0.25
-0.50
-0.75
-1.00

x/B

1.0
0.5

-1.0

-2.0

-2.0

2

3

4

5

G/B

P50

-0.5 -0.4 -0.3 -0.2 -0.1 0.0 0.1 0.2 0.3 0.4 0.5

x/B

6

7

8

9

P50 at D
P50 at S

P34 at D
P34 at S

-1.0
-1.5

1

P34

-0.2

-0.5

-1.5
0

-0.3

0.0

17

1

-0.2
-0.4

0.5

P17 at D
P17 at S

Cp

-0.5

0.0
-0.1

1.0

P1 at DP1 at S

0.0

Cp

-0.2

Cp

-0.4
-0.5
-0.4

(b) 1.25

G/B=0.25
G/B=0.5
G/B=1.0
G/B=1.5
G/B=2.0
G/B=5.0
G/B=10.0
standalone

-0.3

Cp

1.00
0.75
0.50
0.25
0.00
-0.25
-0.50
-0.75
-1.00

10

34
50
0

1

2

3

4

5

G/B

6

7

8

9

10

Fig. 3. Average air pressure coefficient on railway girder: (a) upper surface; (b) upper surface

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3.2.
Flow field diagram
Fig. 4 shows the vorticity diagrams of a standalone girder when facing the wind and the vorticity
diagram when the highway girder is against the wind at different spacing. The wind pressure
peaks all appear at the corner position of the girders, and the separation and reattachment of theflow at the corner of the girders and the impact at the gap are the reasons for the wind pressure
peaks. When D/B <2.0, the downstream girder is completely wrapped in the shear layer formedby the upstream highway girder, becoming a single-flow state, with regular vortex shedding at
the end, leading to more significant aerodynamic interference. The wind pressure of the uppersurface wind fairing increases with the increase of spacing, and the wind pressure of the lower
surface wind fairing increases and then decreases; when D/B =5.0, the upstream wake graduallydissipates in the gap, and the flow pattern of both gradually converges to that of the independent
girder, thus the aerodynamic disturbance gradually decreases.
Standalone highway girder

Standalone railway girder

G/B=0.25

G/B=0.5

G/B=1.0

G/B=1.5

G/B=2.0

G/B=5.0

G/B=10Fig. 4. Vorticity magnitude
4. Conclusions
In this study, when the spacing D/B<2.0, the effect of aerodynamic disturbance on the averagewind pressure coefficient of both upstream and downstream is more significant, and theinterference on the downstream girder is more obvious than that of the upstream. When D/B=5,
the effect of aerodynamic interference is weakened, and the wind pressure of the double-width
girder tends to be similar to that of the standalone girder.
5. References
Laima S.J., Li H., 2015. Effects of gap width on flow motions around twin-box girders andvortex-induced vibrations. Journal of Wind Engineering and Industrial Aerodynamics. 139, 3749.
He X.H., Kang X.M., et al, 2021. Numerical investigation of flow structures and aerodynamic
interference around stationary parallel box girders. Journal of Wind Engineering and Industrial
Aerodynamics, 215, 104610.

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MonitoringBridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

A Trust-Based Approach for Resilient UAV Networks Monitoring Bridges
Francesco Buccafurri1* and Francesca Scoleri2
1
2: University of Reggio Calabria, Italy; email: bucca@unirc.it: Master Graduate at University of Reggio Calabria, Italy; email: scolerifrancesca1@gmail.com*: corresponding author

Abstract: Deterioration of bridges is a very serious issue for transportation agencies andgovernments, not only for the reduction of functionality and reliability of the infrastructure but
also because it can put public safety at risk. The use of multiple Unmanned Aerial Vehicles(UAVs), organized into a multi-hop wireless delay-tolerant self-organizing network, allows us toadopt comprehensive, inexpensive, and fast nondestructive inspections of bridges, thus enabling
effective monitoring of the infrastructure. Unfortunately, the open architecture and the dynamic
topology of the UAV network, makes it vulnerable to a variety of cyber attacks. Among otherpossible attacks, we focus on two types of attacks in this paper. The first regards the case inwhich the compromised UAV tries to send false observations to the base station. The second
happens when the attacker takes control of the UAV and moves it to unexpected trajectories.
Both cases can result in very negative effects. Taking advantage of the specific topology ofbridges and, thus, the way in which the coverage path planning can be organized, we propose, inthis paper, a trust-based mechanism that aims to degrade the reputation of an UAV that attempts
to misbehave. The trust-based mechanism is distributed among UAVs, in such a way that each
UAV can play the role of observer or referee. In the latter case, one UAV is in charge of judging
the behavior of another UAV that is nearby. Coverage path planning is designed so that UAVs
meet frequently during the flight. Using this approach, two global reputation rates for each UAV
are calculated by the base station, based on past interactions. The first is called Content
Reputation Rate (CRR) and refers to the reliability of the content of the observations sent by the
UAVs to the base station. The second is called Position Reputation Rate (PRR) and refers to thereliability of the trajectory traveled by the UAVs. Some cryptographic mechanisms are also
adopted to make the reputation system robust. In addition, slandering and self-promotion attacks
are countered. The outcomes of this work seem promising, because the approach followed may
increase the resilience of UAV networks without requiring additional resources, instead
exploiting the normal task of UAVs, with a small addition of redundancy, which is still desirable
in a critical application setting such as bridge monitoring.
1. Introduction
Unmanned Aerial Vehicles (UAVs) are widely used in the field of civil engineering andinfrastructure monitoring. This is due to the fact that they represent a very flexible tool toaccomplish tasks that, otherwise, would be much more expensive and complex to perform.
Consider that the economic costs of modern UAVs is progressively decreasing, and, moreover,their mobility allows us to implement effective monitoring tasks without infrastructuralinvestments. Finally, they allow visual inspections also in the case of areas hardly accessible forhumans, with no safety risks for the operators, and with the possibility of reuse for multiple

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missions (this is not the case of sensor-based monitoring). Anyway, UVAs and sensors can bealso profitably combined, for example to make easier data collections and sensor control and tosave the energy consumption of wireless sensors. The above considerations are particularly truein the case of road infrastructures such as bridges (especially in the case of big bridges), due totheir conformation and location that makes difficult traditional monitoring.
To reduce the overall mission time and to increase resilience, UAVs often operate as acollaborative group. Usually, the collaboration does not regard the subject of the mission, but theway in which UAVs can communicate with the base station (which is the ground segment of themulti-UAV system). Indeed, a multi-hop wireless delay-tolerant self-organizing network isestablished between the UAVs to route data, according to the paradigm of ad-hoc networks. In
words, routing is collaborative ins the sense that the route depends on when UAVs meet eachother during the flight, in such a way that a one-hop machine-to-machine communication can be
performed. This way, data are forwarded from an UAV to another UAV, until they reach thebase station (and vice versa). Therefore, delays can be generated (thus the network is said delaytolerant). But for the particular application, we do not need real-time routing.
It is rather intuitive that a similar system, being based on the mutual collaboration of UAVs,
strongly relies on the assumption that UVAs do not behave maliciously. However, thisassumption is too strong in the real life, by considering that UAVs perform machine-to-machinecommunication protocols, thus they expose open communication ports, and then are prone tocyber-attacks. Therefore, attackers can corrupt nodes in the UAV network, to make different
attacks, such as the black/gray hole attack, sewage attack, witch attack, etc. This way, not only
data can be eavesdropped, but also message can be tampered or deleted and the whole behaviorof the UAV can be completely compromised.
In this paper, among other possible attacks, we focus on two types of attacks. The first attack
regards the case in which the compromised UAV tries to send false observations to the basestation. The second attack happens when the attacker takes control of the UAV and moves it to
unexpected trajectories. Observe that, both cases can result in very negative effects for the wholesystem and, directly or indirectly, can also put public safety at risk.
Taking advantage of the specific topology of bridges and, thus, the way in which the coveragepath planning can be organized, we propose, in this paper, a trust-based mechanism that aims todegrade the reputation of an UAV that attempts to misbehave.
The contribution of this paper is then to increase the resilience of UAV networks without
requiring additional resources. Indeed, our solution basically exploits the normal task of UAVs,
with a small addition of redundancy, which is still desirable in a critical application setting suchas bridge monitoring.
The structure of the paper is the following. In Section 2, we briefly illustrate the related literature.
In Section 3, we present our trust-based method, while in Section 4 we draw our conclusions.
2. Related WorkIn the scientific literature, there is an increasing interest in the application of UAVs to the field ofcivil engineering (Greenwood et. al, 2019). Non-destructive analysis conducted with UAVs istaking a prominent role in bridge monitoring in the recent years (Feroz and Abu Dabous, 2021;
Duque et. al, 2018). UAVs are organized into fleets in which the UAVs themselves are nodes ofa communication network (Khawaja et. al., 2017). As communication ports of UAVs must be
enabled, cyber-attacks are possible, seriously threaten the reliability of UAV-based monitoring(Krishna and Murphy, 2017; Khan et. al., 2020). Many different approaches are adopted to98Bridge Engineering Institute Conference 2023 (BEI-2023)
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protect UAV networks. A first important objective is to prevent impersonation attacks. This is
typically obtained by using classical cryptographic-based mechanism (Asokan et. al., 2007).
Additional techniques can be used to counter other possible attacks. Our paper is framed in thecontext of trust-based approaches. The work that appears most related to our paper is (Ge et. al,2020). In this work, the authors propose a provenance-aware distributed trust model for UAVNs
that aims to achieve accurate peer-to-peer trust assessment and maximize the delivery of correct
messages received by destination nodes. Our paper also adopts a trust-based approach, but
focuses on different security issues, as illustrated in the introduction.
3. The Proposed Method
In this section, we describe the trust-based mechanism we propose to protect an UAV network
used for bridge monitoring.
The trust-based mechanism is distributed among UAVs, such that each UAV can play the role ofobserver or referee. In the latter case, one UAV is in charge of judging the behavior of another
UAV that is nearby. It is worth noting that the solution is specifically designed for the case ofbridges, among all possible other applications. Indeed, as mentioned earlier, coverage path
planning exploits the specific shape of bridges to ensure that UAVs meet frequently during theflight. Using this approach, two global reputation rates for each UAV are calculated by the basestation, based on past interactions of UAVs. The first is called Content Reputation Rate (CRR)
and refers to the reliability of the content of the observations sent by the UAVs to the basestation. The second is called Position Reputation Rate (PRR) and refers to the reliability of the
trajectory traveled by the UAVs. Some cryptographic mechanisms are also adopted to make thereputation system robust against impersonation.
Let us see more in detail the approach. The first step of our approach is the definition of a bridgespecific coverage path planning, which, based on the particular topology of the region to bemonitored (a region that extends longitudinally), allows us to define a coverage plan that ensuresthat UAVs meet in flight so that they can communicate with high frequency. For space reason
we cannot expand here the definition of the method, but it can be simply based on the applicationof any already known technique to a virtual definition of bridge regions, defined by overlapping
squares that cover the entire monitoring surface with some addition shifting square used for
shuffling UAVs.
Once this step is performed, we can define how to set the trust-based mechanism.
Recall that, according to the communication protocol of the UAV network, when one UAV
encounters another UAV it sends the observation (photo, thermoscan, or other evidence) to theneighbor, so that it will then be conveyed to the base station. Thus, the fact that UAVs
periodically must meet each other is native in the communication protocol. However, the bridgespecific coverage path planning, is used to make this feature more robust.
We can deal with two types of attacks:(i) the case in which the compromised UAV sends false
observations, and (ii) the case in which the compromised UAV takes a physical path different
from the predetermined one, which may mean that the attacker has taken control of the piloting.
When the UAV A sends the observation Ob to UAV B, it also sends its GPS location P, so it
sends the tuple (ID, Ob, P), where ID is its identity – we adopt cryptographic mechanisms togive assurance to the identity. The UAV B, which receives the observation, being nearby, canperform the same kind of observation, whose type, position etc., are all included in Ob, orotherwise can perform some checks (possibly relying on an artificial intelligence engine).
Thanks to this verification, when forwarding the message, it adds a score to that message, called99

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content reputation rate - CRR. Similar thing it does about the position, because it can assess
whether or not the GPS coordinates that UAV A is communicating to it are plausible, since it cancompare them with its own. This way, the referee can compute another reputation rate, called
position reputation rate (PRR). Thus, B will forward the message: (ID, Ob, P, CRR, PRR), which
will eventually reach the base station.
Based on this reputation model, the base station receiving the messages will detect compromised
UAVs, because if an UAV is misbehaving, then it will receive negative rates from several
neighboring UAVs, which play the role of referee. It is easy to see that self-promoting attacks
are not possible by design (since the provided level of assurance of the UAV identity prevent
impersonation attacks and thus an UAV cannot simulate to be a referee). Moreover, slandering
attacks are contrasted by the fact that the computed reputation is global, in the sense that itderives from interactions with different UAVs. Therefore, the effect of a possible fake referee, isneutralized.
4. Conclusions
In this paper, a trust-based distributed model is proposed to counter two types of cyber-attacks
that can threaten UAV networks used for bridge monitoring. The proposed approach is based on
the concept of reputation, implemented via two global reputation rates, one related to thereliability of observation contents, the other related to the reliability of the UAV position. The
outcomes of the paper appear promising, because the solution aims to increase the resilience of
UAV networks, which is a very relevant aspect in the considered application scenario.
5. References
Asokan, N., Kostiainen, K., Ginzboorg, P., Ott, J., and Luo, C. 2007. Applicability of identitybased cryptography for disruption-tolerant networking. In Proceedings of the 1st international
MobiSys workshop on Mobile opportunistic networking (pp. 52-56).
Duque, L., Seo, J., and Wacker, J. 2018. Bridge deterioration quantification protocol using UAV.
Journal of Bridge Engineering, 23(10), 04018080.
Ge, C., Zhou, L., Hancke, G. P., and Su, C. 2020. A provenance-aware distributed trust model
for resilient unmanned aerial vehicle networks. IEEE Internet of Things Journal, 8(16), 1248112489.
Feroz, S., and Abu Dabous, S. 2021. Uav-based remote sensing applications for bridge conditionassessment. Remote Sensing, 13(9), 1809.
Greenwood, W. W., Lynch, J. P., and Zekkos, D. 2019. Applications of UAVs in civil
infrastructure. Journal of infrastructure systems, 25(2), 04019002.
Khan, N. A., Brohi, S. N., and Jhanjhi, N. Z. 2020. UAV’s applications, architecture, security
issues and attack scenarios: A survey. In Intelligent Computing and Innovation on Data Science:
Proceedings of ICTIDS 2019 (pp. 753-760). Springer Singapore.
Khawaja, W., Ozdemir, O., and Guvenc, I. 2017. UAV air-to-ground channel characterization
for mmWave systems. In 2017 IEEE 86th Vehicular Technology Conference (VTC-Fall) (pp. 15). IEEE.
Krishna, C. L., and Murphy, R. R. 2017. A review on cybersecurity vulnerabilities for unmanned
aerial vehicles. In 2017 IEEE International Symposium on Safety, Security and Rescue Robotics(SSRR) (pp. 194-199). IEEE.

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Evaluation Method for Grouting Condition in Post-tensioned PC Bridges
Considering Directive Propagation Properties of Elastic Waves in the WideRange Ultrasonic Testing (WUT)
Takanori Kinoshita1, Yoshino Sako2, Kuniharu Fukushima3, and Isamu Yoshitake4*
1: H and B System Co, Ltd., Tokyo, Japan; email: kinoshita@hbsys.co.jp: Nippon PS Co. Ltd., Fukui, Japan; email: yoshino.sako@nipponps.co.jp
3: Nippon PS Co. Ltd., Fukuoka, Japan; email: k.fukushima@nipponps.co.jp
4: Yamaguchi University, Yamaguchi, Japan; email: yositake@yamaguchi-u.ac.jp
2*: corresponding author

Keywords: Ultrasonic testing (UT) ; Prestressed concrete ; Elastic wave ; Laser Doppler
vibrometer (LDV) ;Non-destructive inspection (NDI)
Abstract: Wide-range ultrasonic testing (WUT) is occasionally used to examine the grout
filling-condition of post-tensioned PC bridges. The study focuses on improvement of inspection
method of grouting condition. For this purpose, laboratory and field tests were conducted in the
study. In the fundamental test, semicylinder specimens made with concrete and mortar wereprepared to investigate the directivity of elastic waves. A laser Doppler vibrometer was used as
the alternative receiver of WUT. The test results revealed that the elastic-waves propagation in
mortar was highly linear. In addition, it was observed that the coarse aggregate influenced on the
directivity of propagation waves. The elastic waves in concrete were in the range of -30o to 30oapproximately. Based on the fundamental test, grouting conditions in the duct of actual PC
bridges were examined to investigate the applicability of the improved system. It was confirmedthat the developed inspection method can improve the accuracy for grouting conditions.
1. Introduction
Wide-range ultrasonic testing (WUT) is a non-destructive inspection method to evaluate grout
filling-condition of post-tensioned PC bridges. The WUT evaluates the complete or incomplete
grouting by comparing the peak frequency of elastic waves reflected from the duct [Fukushima
et al. 2018]. The thresholding method cannot determine the grout-filling condition when multiple
peaks are observed in the WUT [Hamaoka et al. 2014]. An additional evaluation method isrequired for improving the reliability of the WUT inspection. To examine the propagation of thetransmitted waves in the WUT, fundamental test was conducted by using semi-cylindricalconcrete specimens. The study focused on the time-variation of elastic waves to evaluate thegrout filling condition. The purpose of the study was to propose an additional evaluation method
based on the propagation characteristics of elastic waves in concrete. The applicability of the
proposed method was confirmed at a post-tensioned PC bridge that embedded void and groutfilled ducts. This paper reports the fundamental experiment and verification of the proposed
evaluation for the WUT.

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2. Fundamental Test
In the fundamental test, semi-cylindrical specimens made with concrete and no-shrink mortar
were used. Dimensions of the specimen were 400 mm diameter and 400 mm high. Figure 1(a)shows an overview of the specimen. The transmitting transducer was fixed on the center of theplane. To observe the elastic waves, a laser Doppler vibrometer (LDV) was used in the test.
Figure 1(b) shows the measurement system.
21 points measurement at intevals of 15.7mm(5degree equivalent)measured signal

15.7@20=314

trigger signal

Oscilloscope

directivity angle

LDV(Receiver)
Pulser-receiverSpecimen
voltage pulse

400

Fix

400

Ultrasonic transducer(Transmitter):Transmitter (Ultrasonic transducer):Reciever(LDV)(a)(b)
Fig. 1. Fundamental test set-up: (a) overview of test specimen; (b) measurement system
The relationship between directivity angle and the maximum amplitude (absolute value) of theobserved elastic waves is shown in Figure 2.
2
Amplitude (V)Concrete
1.5

Mortar1
0.5
0-50° -40° -30° -20° -10° 0° 10° 20° 30° 40° 50°
Directivity angle(°)

Fig. 2. Maximum amplitude of received wave

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The amplitude of the mortar specimen reached the maximum value of 1.9 V at a directivity angleof 0o in the range of the directivity angle of ±25o. The result shows that the directivity(straightness) of the transmitted wave is remarkable in mortar. On the other hand, for theconcrete specimens, the amplitude was approximately 0.4 V in the range of -10° to 15°
directivity angle. The maximum amplitude in concrete was lower than 25% of that in mortar.
The observation was caused from the coarse aggregate effect, the scattering and attenuation oftransmitted waves. The amplitude out of the range of -30° to 25° in concrete was almost constant,the elastic waves in concrete were in the range of -30° to 30° approximately. The experimental
results imply that the closer arrangement of transducers may be effective because of the highdirectivity of the transmitted wave by the WUT transducer.
3. Field Test
In the field test, WUT inspection was conducted on post-tensioned PC bridges underconstruction. Based on the fundamental experiment, the transducers were placed along the ductswith a minimum distance of 97 mm between them. For comparison, the experimental parameters,
the transducers distance of 150 / 200 / 500 mm, were also added. Note is that the transducers
spacing of 200, 500 mm is often employed in actual surveys of the WUT. Figure 3 illustrates
time-variation of the waveform in the WUT measurement with the transducer distance of 97 mm.
30000

Empty duct
Grout-filled duct

Amplitude (mV)

20000
10000
0
-10000-20000-30000Refleted wave arrival time tr
0

10

20

30

40
50
60
Time (10-6s)

70

80

90

100

Fig. 3. Example of waveforms (distance between transducers is 97mm)
The amplitude of the waveform measured in concrete embedding the empty duct was greaterthan the grout-filled duct. Based on the characteristics, the study focused on the amplitude of the
waveform that measured during the time tr defined in Equation (1).
(1)
where, a is distance between transducers (center to center), cd is covering depth of duct, φ is duct
diameter, and ν is elastic wave velocity in concrete.
To evaluate the grout filling condition, the averaged amplitude |Vav| was used in the study. The
definition of |Vav| is described in Equation (2). The averaged amplitude |Vav| in this study isshown in Equation (2). Figure 4 shows the averaged amplitude |Vav| in the field test.

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Distance between transducers (mm)

(2)

33
52

500

Grout-filled duct
Empty duct
1518
1621

200

2862

150

4042

3002

97
0

2000
4000
Averaged amplitude (mV)

5152
6000

Fig. 4. Averaged amplitude before and after grouting (a=97~500)
Figure 4 confirms that the amplitude of the empty duct was greater than the grout-filled duct. In
the transducers distance a of 97 mm, the averaged amplitude |Vav| indicated a significant
difference in empty and grout-filled ducts. The observed |Vav| at the empty duct was 5152 mV,
which was around 1.7 times greater than the |Vav| of 2418 mV in the grout-filled duct. It was
confirmed that the grout filling condition can be determined from the differences of averaged
amplitude |Vav|.
4. Conclusions
A fundamental experiment was conducted to examine the directivity of the WUT transducer
using half-cylinder specimens. Afterward, an evaluation of the grout filling conditions was
attempted on the PC bridge under construction by WUT measurements based on the propagation
characteristic. The main findings in this study are as follows:- The significant directivity of elastic waves from the WUT transducer was confirmed.
- The averaged amplitude from the unfilled duct was approximately 1.7 times higher than theamplitude from the filled duct.
- The averaged amplitude of reflected waves is useful for evaluating the grout filling conditionin PC duct.
5. References
1. Fukushima, K., Amaya, K., Kinoshita, T., Hara, M., Yamada, K., and Yoshitake, I., Field
test of Wide-Range-ultrasonic Testing (WUT) to detect incomplete grout in posttensioned
prestressed concrete bridges, 2018, The International Federation for Structural Concrete 5thInternational fib Congress, Melbourne, Australia.
2. Hamaoka, K.; Aoki, K., Hara, M., and Kinoshita, T., The filling survey of PC grout using
the wide range ultrasonic testing, Journal of Prestressed Concrete, Japan, 2014, Vol.56,
No.6, pp.35-40. (in Japanese)104Bridge Engineering Institute Conference 2023 (BEI-2023)
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Detection of Defects in Prestressed Concrete Bridge Box
Girders Using GPR Image Reconstruction
Wael Zatar1, Hien Nghiem2, and Hai Nguyen3
1 Professor and Director, Encova Center for Engineering and Safety, College of Engineering andComputer Sciences, Marshall University, Huntington, West Virginia, United States ofAmerica; email: zatar@marshall.edu
2 Post-Doctoral researcher, Marshall University Research Corporation, Huntington, West
Virginia, United States of America; email: nghiem@marshall.edu
3 Scientist, Engineer Research and Development Center, Urbana-Champaign, Illinois, UnitedStates of America; email: nguyenhai@marshall.edu*: corresponding author

Keywords: nondestructive evaluation; ground-penetrating radar; synthetic aperture focusing
technique; image reconstruction; 3D and 3D visualization; reinforced-concrete structures.
Abstract: Forty two percent of all bridges in the United States are at least 50 years old, and 46,154,
or 7.5 percent of, bridges are considered structurally deficient. Corrosion of reinforcing steel andprestressing strands is a major cause of deterioration, reduced durability, or failure of bridge
structures. Visual inspection typically examines surface discontinuities and relies heavily on
inspectors’ subjective assessments. Ground penetrating radar (GPR) provides a robust nondestructive testing method that has gained acceptance for evaluating subsurface condition of
infrastructure components. This research effort aims at detecting defects in reinforced-concrete
bridge decks using GPR systems. An enhanced synthetic aperture focusing technique was
employed to reconstruct two- and three-dimensional GPR images. A method to determine timezero is developed to improve the quality of the reconstructed images. The method was validated
by evaluating reinforced-concrete slab specimens with internal defects. The validated technology
is used to non-destructively evaluate full-scale prestressed concrete bridge box girders that haverecenetly been decommisioned from the CR 30 Bridge over Tom Creek in West Virginia. Thesebox girders were reported to be in fair condition during the lastest routine inspection, completed
in accordance with the National Bridge Inspection Standards. The proposed GPR visualization
protocol was found to provide a great promise in detecting internal reiforcements and defects ofthe prestressed concrete bridge box girders. This research will benefit West Virginia Departmentof Transportation as the agency continues to safely maintain and preserve its 7,291 bridges.
1. IntroductionMore than 40 percent of all bridges in the United States are at least 50 years old and 7.5 percent,or 46,154, of the nation’s bridges are considered structurally deficient. Corrosion of reinforcing
steel and prestressing strands is a major cause of deterioration, reduced durability, or failure ofbridge structures. Roughly, twenty six percent of reinforced and prestressed concrete highway
bridges in the United States need repair or replacement. A recent estimate for the nation’s backlog

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of bridge repair needs is $125 billion. The United States needs to increase spending on bridge
rehabilitation from $14.4 billion annually to $22.7 billion annually, or by 58%, to improve thecondition. At the rate of investment provided in the past ten years, it will take until 2071 to makethe repairs that are currently necessary, and the additional deterioration over the next 50 years will
become overwhelming (Report Card for America’s Infrastructure, 2021). The nation needs asystematic program for bridge preservation whereby existing deterioration is recognized whiledeveloping preventive maintenance protocols. West Virginia Department of Highways (WVDOH)
maintains 7,291 bridges. Twenty one percent, or 1,531, of these WVDOH bridges are structurally
deficient. WVDOH reported an increase of structurally deficient bridges by 522 in the last 5 years.
Moreover, WVDOH maintains 40.7 million square feet of bridge decks, of which approximately
6.6 million square feet is structurally deficient (Report Card for America’s Infrastructure, 2021).
Heavily relying on subjective assessment, bridge visual inspection typically examines surface
discontinuities of its components. Geophysicists have used GPR in soil explorations for the pastfew decades. The GPR technology has recently gained acceptance in evaluating subsurface
condition of infrastructure components (ACI Committee 228, 2013). The technique showed
potential in both detecting rebars and defects, and in estimating concrete cover and thickness ofbridge decks (Alani et al., 2013, Benedetto et al., 2012, and Garg, S. and Misra, 2020). The primaryobjective of this research is to fill a knowledge gap in GPR imaging of reinforced and prestressed
concrete transportation structures. The study aims at detecting defects in reinforced-concrete
bridge decks and prestressed concrete box girders using GPR systems. Two- and three-dimensional
visualization of concrete structures will assist determining which areas need further attention, or
if components or the entire bridge should be replaced.
2. GPR System and Synthetic Aperture Focusing Technique
non-destructive testing electromagnetic energy GPR waves are subjected to attenuation and loss ofenergy as the waves propagate in a material. In materials with low conductivity such as dry sand,ice, and dry concrete, signals stay immaculate longer and are, thus, able to penetrate a considerable
depth into the material. In contrast, GPR energy will be absorbed before getting far in conductive
materials such as saltwater and wet concrete. GPR testing, therefore, is suitable for inspecting
construction materials such as concrete, sand, wood, and asphalt. A typical GPR antenna consists ofa transmitter (T) and receiver (R), one transmits the signal, and the other receives the reflected signal(Figure 1). GPR with 1.6 GHz antenna and 2.3” (58 mm) T-R offset was used in this study. Signal
amplitudes vary from minimum negative to maximum positive corresponding to the color
spectrum ranging from black to white in gray-color scale and from dark-blue to dark-red in jetGPR
Control
Unit
Antenna Cable

T-R offset(58 mm)TX

RX

Ground-coupled
Antenna 1.6GHz

Concrete SlabRebar

Fig. 1. Principle of GPR testing method

Fig. 2. Raw images of a B-scans
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color scale (Figure 2). Long defects are displayed in localized black stripes (Figure 2a) or darkblue stripes (Figure 2b) for negative signal and V-shaped white stripes (Figure 2a) or yellow-red
stripes (Figure 2b) for positive signal. Images obtained from GPR device’s proprietary software
might not result in accurate identification of rebars and defects.
3. Enhanced GPR Imaging and Visualization
Since the Synthetic Aperture Focusing Technique (SAFT) focuses signal amplitudes at points inthe sampling area of large scans by coherent superposition, its utilization may result in improving
image quality of internal rebars and voids (Zatar et al., 2020 and 2021). An enhanced synthetic
aperture focusing technique (SAFT) was employed to reconstruct two- and three-dimensional GPR
images. A correction function was developed to adjust reflection locations of the defects. A method
to determine time-zero was developed to improve the quality of the reconstructed images.
Computer code with a fully graphical interface, written in Delphi, was developed to overcome the
GPR device commercial post-processing software’s shortcoming in reconstructing images.
4. GPR Imaging and Visualization of Prestressed Concrete Bridge Box Girders
The validated GPR technology is employed to non-destructively evaluate full-scale prestressed
concrete bridge box girders that have recently been decommissioned from the CR 30 Bridge over
Tom Creek in West Virginia. The box girders were reported to be in fair condition during thelastest routine inspection that is completed in accordance with the National Bridge Inspection
Standards. Figure 3 shows a longitudinal section of the box girders. Each girder is 32.5’ long. The
girders are 36” in width and 17” in height. Each girder has two 10” diameter circular voids (Figure4). Eleven tendons were placed near the bottom of each girder, two groups of five tendons with 2”
spacing on the sides and one tendon at the center. Four #4 rebars were placed at the top of thegirders. Stirrups with 12” spacing were used.
GPR data was collected, 97 scan lines in the lateral direction and 9 scan lines in the longitudinal
direction, with scan spacings of 4”. All the bottom 11 prestressing tendons were clearly detectedin the constructed image from B-scan in Figure 5. Reflection images showed two red strips because
the 5 tendons on each side are closely arranged. Figure 6 shows 2D reconstructed images obtained
from GPR top scans of one girder. They were constructed at 20”, representing the solid part of thecross section, and at 80”, representing the part of the cross section with two voids, from end of one
box girder of CR 30 Bridge over Tom Creek in West Virginia. It can be observed that rebars andvoids were detected at 20” and 80” distances from the girder’s end. Figure 7 shows B-scan in thelongitudinal direction where the spacing of stirrups can be obtained accurately. This
reconstructed image also shows that the concrete cover for rebars is uneven.
11'6"

2'

9'6"

32'6"

11'6"

2'

Fig. 3. Longitudinal section of prestressed concrete box girders of CR 30 Bridge over Tom Creek

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Ø10

Ø10
4#4
rebars17

Stirrup12" spacing

Te
Tendons
36

Fig. 4. Cross section of prestressed concrete box girders of CR 30 Bridge over Tom Creek, WV

Void

Void
Tendons

TendonTendonsFig. 5. 2D reconstructed image obtained from GPR bottom scans

Rebars(a)RebarsVoids(b)Fig. 6. 2D reconstructed images obtained from GPR top scans: (a) at 20” from end of box girder;(b) at 80” from end of box girder of CR 30 Bridge over Tom Creek in West Virginia

Stirrups

Main rebars

Fig. 7. 2D reconstructed images along CR 30 Bridge box girders (obtained from GPR top scan)

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5. Conclusions
Enhanced synthetic aperture focusing technique and ground-penetrating radars were employed toimage and assess current condition of reinforced and prestressed concrete bridges. The developed
method and software were validated by non-destructively testing prestressed concrete box girders
decommisioned from CR 30 Bridge over Tom Creek in West Virginia. The GPR reconstructed
images clearly visualize rebars, tendons, and voids of the box girders. Moreover, the images
showed stirrup spacings and rebar locations along the prestressed concrete bridge box girders. This
research will benefit West Virginia Department of Transportation as the agency continues to safely
maintain and preserve the 7,291 bridges included in its bridge inventory.
6. Acknowledgment
Special acknowledgement is due to Mr. Jimmy Wriston, West Virginia Secretary of Transportation
and Commissioner of Highways, for the superior technical guidance and thoughtful monitoring of
this project over the past several years. The authors would also like to acknowledge the financial
contribution provided by the West Virginia Division of Highways to carry out this study. The
authors would like to thank Dr. Cumhur Cosgun, Dr. Tu Nguyen, and Dr. Kien Dinh for the earlier
developmental NDT research at Encova Center for Engineering and Safety at Marshall University.
7. References
Report Card for America’s Infrastructure. 2021. American Society of Civil Engineers,
www.infrastructureportcard.org
ACI Committee 228. 2013. Report on Nondestructive Test Methods for Evaluation of Concrete in
Structures, Report ACI 228.2R-13, American Concrete Institute.
Alani, A.M., Aboutalebi, M. and Kilic, G. 2013. Applications of ground penetrating radar (GPR)in bridge deck monitoring and assessment, Journal of Applied Geophysics, 97, pp.45-54.
Benedetto, A., Manacorda, G., Simi, A. and Tosti, F. 2012. Novel perspectives in bridges
inspection using GPR, Nondestructive Testing and Evaluation, 27(3), pp.239-251.
Garg, S. and Misra, S. 2020. Efficiency of NDT techniques to detect voids in grouted posttensioned concrete ducts, Nondestructive Testing and Evaluation, pp.1-22.
Zatar, W.A., Nguyen, H.D. and Nghiem, H.M. 2020. Ultrasonic pitch and catch technique for nondestructive testing of reinforced concrete slabs, Journal of Infrastructure Preservation andResilience, 1(1), pp.1-13.
Zatar, W.A., Nguyen, H.D., Nghiem, H.M. and Cosgun, C. 2021. Non-Destructive Testing of
GFRP-Wrapped Reinforced-Concrete Slabs, Transportation Research Board 100th Annual
Meeting, paper No. TRBAM-21-03354.

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Prestressing Force Distribution in Post-Tensioned
I-Girder Bridges Measured Using Smart Strands
Se-Jin Jeon1*, Sang-Hyun Kim2, Sung Yong Park3, and Sung Tae Kim4
1

: Department of Civil Systems Engineering, Ajou University, Suwon-si, Korea; email:conc@ajou.ac.kr
2: Convergence Technology Research Team, Daewoo Institute of Construction Technology,
Suwon-si, Korea; email: sanghyun.kim@daewooenc.com
3: Department of Structural Engineering Research, Korea Institute of Civil Engineering andBuilding Technology, Goyang-si, Korea; email: sypark@kict.re.kr
4: Department of Structural Engineering Research, Korea Institute of Civil Engineering andBuilding Technology, Goyang-si, Korea; email: esper009@kict.re.kr*: corresponding author

Keywords: prestressed concrete; prestressing tendon; strand; prestressing force; prestress loss;
fiber optic sensor; fiber Bragg grating
Abstract: The proper distribution of prestressing force(PF) is the basis for the design ofprestressed concrete(PSC) structures. However, the PF distribution obtained by predictive
equations of prestress losses has not been sufficiently validated by comparison with measured
data due to the poor reliability and durability of conventional sensing technologies. Therefore,
the Smart Strand with embedded fiber optic sensors was developed and applied to PSC structures
to investigate the short- and long-term characteristics of PF distribution as affected by friction,anchorage seating, elastic shortening, concrete creep and shrinkage. The data measured in a 20
m-long full-scale specimen and a 60 m-long PSC girder bridge were analyzed by comparing
them with the theoretical estimation obtained from several design equations. The study results
also confirmed the need to compensate for the temperature variation in the long-term monitoring
to derive the actual mechanical strain related to the PF. We expect our developed Smart Strand tobe applied practically in PF measurement for the reasonable safety assessment and maintenanceof PSC structures by improving several of the existing drawbacks of conventional sensors.
1. Introduction
Determining the proper distribution of the prestressing force(PF) in the tendons is crucial in thedesign of prestressed concrete(PSC) structures because it largely affects safety and serviceability.
The PF varies along the length of a tendon and over time due to the short- and long-term losses
of prestress. Although the PF distribution is usually estimated using several predictive equations
for the prestress losses during the design of a PSC structure, these equations can only
approximate the PF distribution. Although many attempts have been made to validate thetheoretical distribution of the PF by using conventional sensing technologies, their outcomes
were not very successful due to several drawbacks that led to unreliable data.

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In order to overcome the abovementioned drawbacks and limitations of existing sensing
technologies in the estimation of the PF distribution, Smart Strands with embedded fiber optic
sensors have been developed recently(Kim et al., 2020; Kim et al., 2022).
The dimensions of the Smart Strand are almost identical to those of a regular seven-wire strand
widely used in PSC structures, as shown in Fig. 1. The steel core wire of the regular strand is
replaced with carbon fiber reinforced polymer(CFRP) and a fiber optic sensor with several
embedded FBGs, while the CFRP core wire is manufactured.

5.3 mm15.3 mm

Core wire(CFRP)

Protective
coverCore wire(CFRP)
Helical wire(steel)Helical wire(steel)0.9 mmFiber optic sensor(FBG type)

Fiber optic sensor(FBG type)5.0 mmFig. 1. Configuration of a Smart Strand
The wavelength of a reflected light wave measured at each FBG can be converted to the strain of
Smart Strand by Eq. 1 when the effect of temperature is compensated(Jeon et al., 2022).
ε=
1  ∆λ− α f + ξ + (1 − pe ) (α smart − α f )  ∆T 

1 − pe  λB
(1)

where ε : mechanical strain of Smart Strand, pe : photo-elastic coefficient(=0.22), ∆λ = λ − λB :wavelength shift, λ : measured wavelength, λB : base wavelength, α f : thermal expansion
coefficient of an FBG(=0.5×10−6 /°C), ξ : thermo-optic coefficient(=6.2×10−6 /°C), α smart :thermal expansion coefficient of Smart Strand(=8.68×10−6 /°C), ∆T = T − TB : temperature change,
T : measured temperature, and TB : base temperature. The strain obtained in Eq. 1 can further be
converted to the PF using the force–strain relationship of P = ( E p Ap ) smart ε , where ( E p Ap ) smart :equivalent E p Ap of Smart Strand, and E p and Ap : modulus of elasticity and cross-sectional areaof a strand, respectively.
2. Application of Smart Strands to Post-Tensioned Structures
Fig. 2 shows the post-tensioned 20 m-long full-scale specimen, where three types of Smart
Strand with three, five, and seven equally spaced FBGs, respectively, were fabricated andselectively inserted into each duct, together with regular strands. Fig. 3 shows the PSC girder
bridge with a 60 m span, where one Smart Strand was applied to one of the six ducts arranged in
one of the 10 girders. The Smart Strand had seven FBGs and was inserted into the duct, together
with 11 regular strands.

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Centerline

10,000

300

600

1300
2000

T2
T3
Unit: mm

300

φ85

T1

2@200=400

300

2000

2@700=1400

600

Fig. 2. Post-tensioned full-scale specimen

4000

4000

20,00020,000FBG 1 FBG 2 FBG 3

4000

4000

FBG 5 FBG 6 FBG 7

FBG 4

Fig. 3. PSC girder bridge
3. Short-Term Losses of Prestress
The PF distributions of the full-scale specimen and the PSC girder as affected by short-term
prestress losses were represented in Fig. 4, where it apparently shows how the PF distributions
are affected by each short-term prestress loss. The PF gradually decreases from the live end dueto the friction loss during tensioning. Then, the PF distribution near the anchorage zone
decreases due to the anchorage seating loss. The PF distribution further decreases according tosubsequent tensioning of the remaining strands included in the other ducts. The measured PF
distributions were also compared with those theoretically estimated. The overall trend of PF
distributions in each stage of short-term prestress losses was similar for the measured andtheoretical values, although they showed some differences in absolute values. These results in
Fig. 4 confirmed the usefulness of the existing predictive equations for the short-term prestresslosses in estimating the PF distribution for design purposes. However, the theoretical PF
distribution was sensitive to the assumed values for the friction coefficients and slip
displacement, which highlights the importance of reasonable design parameters.

200

Anchorage
seating loss

measured(f)measured(a)Theoreticaltheory(f)theory(a)Elastic
shortening loss

Measuredmeasured(e)theory(e)Theoretical jacking force

180
160
140
120

0

5

10

15

Prestressing force (kN)

Prestressing force (kN)

Friction
loss
Measured

20

Theoretical

200

Frictionloss

measured(f)Anchorage
seating loss

measured(a)theoretical(f)Elastic
shortening loss

theoretical(a)180theoretical(e)Theoretical jacking force

160
140
120

0

10

20

30

40

Distance (m)Distance (m)(a) Full-scale specimen(b) PSC girder bridge

Fig. 4. Short-term PF distribution in measurement and theory112

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4. Long-Term Losses of PrestressFig. 5(a) presents the PF variation at mid-span of the full-scale specimen over time in bothmeasurement and theory. Three Smart Strands(H1, H7, and H12) of T1 in Fig. 2 wererepresentatively analyzed herein. Zero-day corresponds to the time after tensioning when the
short-term losses had just occurred. The predictive equations for the long-term prestress losses
induced by creep and shrinkage were compared with the measurements: ACI 209R, ACI
423.10R, and Eurocode 2. Among the three equations considered, Eurocode 2 showed relatively
good agreement with the averaged measurement values.
Fig. 5(b) compares PFs at mid-span of the PSC girder between the predictive equations and themeasurement. The figure suggests that several major construction stages caused significant PF
fluctuations, in addition to the decreasing trend induced by long-term prestress losses. Overall,
Eurocode 2 was in good agreement with the measured values, similar to the full-scale specimen.
200

Measured (H7)
Measured(T1-H7)

Measured
Measured(H12)(T1-H12)

Theoretical(ACI-ASCE)
Theoretical(ACI)
Theoretical(ECTheoretical (ACI 209R)
Theoretical (Eurocode
2) 2)

Measured
Measured(Average)(T1-Avg.)

Theoretical (ACI 423.10R)

Prestressing force (kN)

Prestressing force (kN)

Measured
Measured(H1)(T1-H1)

Theoretical jacking force

180
160
140
120

0

50

100

150
200
Time (days)

250

300

350

200

Measured
Measured

Theoretical(ACI)
Theoretical(ACI 209R)

Theoretical(EC2)
Theoretical (Eurocode
2)

Theoretical(EC(Eurocode
2, CS)
Theoretical
2 with construction stage)

180

Theoretical jacking force
Deck placement

160

Secondary prestressing

140
120

Theoretical(ACI-ASCE)
Theoretical(ACI 423.10R)

050

Placement of barrier
and median strip

100 150 200 250 300 350 400 450 500

Time (days)(a) Full-scale specimen(b) PSC girder bridge

Fig. 5. Long-term PF distribution in measurement and theory
5. Conclusions
The Smart Strand with fiber optic sensors of FBG type was developed to investigate the effect ofprestress losses on the prestressing force(PF) distribution.
The PF distributions were affected by the short-term prestress losses in the order of friction loss,
anchorage seating loss, and elastic shortening loss. Overall trend of the measured and theoretical
PF distributions was similar in each stage of the short-term prestress losses.
Regarding the long-term prestress losses, although the two approaches of measurement andtheory produced a similar long-term trend of the PF distribution, the difference between themvaried depending on the design equations. The equation of Eurocode 2 to estimate the long-term
prestress losses showed relatively good agreement with the measurement, when compared to thatof ACI 209R and ACI 423.10R. This can be attributed to the more refined form of the predictive
equation of Eurocode 2 that addresses the time-dependency of the PF.
6. References
Jeon, S.J., Park, S.Y., and Kim, S.T. 2022. Temperature compensation of fiber Bragg grating
sensors in Smart Strand, Sensors, 22(9), 3282.
Kim, S.H., Park, S.Y., and Jeon, S.J. 2020. Long-term characteristics of prestressing force inpost-tensioned structures measured using Smart Strands, Applied Sciences, 10(12), 4084.

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Kim, S.H., Park, S.Y., Kim, S.T., and Jeon, S.J. 2022. Analysis of short-term prestress losses inpost-tensioned structures using Smart Strands, International Journal of Concrete Structures andMaterials, 16(1), 81-95.

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Detection of Cable Section Loss in Cable-Stayed
Bridge with Electromagnetic Sensor
Imjong Kwahk1, Changbin Joh2, Ji-Young Choi3, Kwang-Yeun Park4, and Joo-Hyung Lee5*
1: Korea Institute of Civil Engineering and Building Technology (KICT), Goyang-si and South
Korea; email: kwakim@kict.re.kr
2: KICT, Goyang-si and South Korea; email: cjoh@kict.re.kr
3: KICT, Goyang-si and South Korea; email: legion@kict.re.kr
4: KICT, Goyang-si and South Korea; email: kypark@kict.re.kr
5: KICT, Goyang-si and South Korea; email: leejoohyung@kict.re.kr*: corresponding author

Keywords: Cable section loss; Non-destructive testing; Electromagnetic sensor; bridge
inspection; Total magnetic flux
Abstract: There are many islands in Korea, and several sea bridges, including cable-stayed
bridges, have been constructed and serviced. Since cables of the cable-stayed bridges are a
critical member for the structural safety of bridges, they are included in the subject of regularbridge inspection. However, inspecting its condition is practically impossible since the cable isinside the duct. Although various non-destructive evaluation methods, including Magnetic Flux
Leakage, have been proposed and applied, there are still limitations to being applied in an actual
bridge environment. Therefore, the Korea Institute of Civil Engineering and Building
Technology developed a detachable cylindrical electromagnetic sensor using Faraday's law of
induction, which states that changes in magnetic flux generate force, to detect section loss of thecable located in the duct. The section damage detection principle of the sensor is as follows. First,
wrap the sensor around the cable and install it. When alternating current is applied to the primarycoil of the solenoid-type sensor, a magnetic field is formed in the longitudinal direction of thesensor. At this time, when the sensor is moved, an induced electromotive force is generated
according to Faraday's law. The magnitude of the force is proportional to the number of turns of
the coil and the change rate of the magnetic flux passing through the cross-sectional area
perpendicular to the magnetic field inside the sensor. So, the induced electromotive force is also
decreased when the cable cross-section is reduced. By measuring the force with the secondary
coil located in the sensor's center, it is possible to detect the section loss. Non-destructive tests
with the developed sensor were performed on specimens fabricated to replicate the cable-stayed
bridge's cable, and the damage detection capacity was evaluated. In addition, NDT was
conducted on cables of an actual cable-stayed bridge to evaluate on-site applicability. When the
EM sensor is improved through further research, it is expected that it could be applied as a meansof cable inspection of the in-serviced bridges in Korea.

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1. Introduction
The sea surrounds Korea on three sides, with many islands, especially in the West and South
Seas. So, dozens of cross-sea bridges connect the island to the mainland or the island to theisland. Because it has several advantages: low cost, ease of construction, and aesthetics (Zhaoand Tonias, 2013), the cable-stayed bridge is commonly constructed as a cross-sea bridge in
Korea.
The cable is the primary structural member supporting the superstructure loads for thecable-stayed bridges. Therefore, if severe defects or deterioration occurred in the cable, thebridge's structural safety could be severely compromised. Especially in the case of cross-sea
bridges, cable corrosion due to salt damage could be very harmful to safety. However, despite its
structural importance, there is no method to inspect the inside cable duct. Only it is possible toestimate the condition inside the cable by visual inspection and hammering.
Various non-destructive testing (NDT) methods, such as Magnetic flux leakage andImpact echo, have been suggested and applied to find defects inside the bridge duct. However, it
was confirmed that the workability and accuracy were poor at the actual bridge site (Seoul
Metropolitan Government, 2017). Therefore, it needs to improve the detection performance ofcable section loss for effective bridge maintenance. The Korea Institute of Civil Engineering andBuilding Technology (KICT) developed an electromagnetic (EM) sensor that could detect thesection loss of bridge cable.
2. The Electromagnetic Sensor to Detect Cable Section Loss
The developed EM sensor’s basic detection principle is from the EM induction, commonly
known as Faraday’s law. According to the law, the electromotive force around a closed patheqauls to the negative of the time rate of change of the magnetic flux enclosed by the path
(Jordan and Balmain, 1968).
When alternating current is applied to the solenoid-type sensor surrounding the cable, amagnetic field is formed inside the sensor. Then, the magnetic flux density changes with time, an
induced electromotive force (emf), Vind, is generated according to Faraday’s law. It could be
expressed in Eq. 1 (Kwahk et al., 2020). In here, N is the number of coil turns, Ф is the totalmagnetic flux, B is magnetic flux density, and A is cross-sectional area inside the sensor.
dΦd  Vind =−N=− N ∫ BdA
dt
dt A

(1)

It is assumed that non-magnetic materials and air inside the sensor do not affect the induced
electromotive force. Then, if the cable cross-section reduction has occurred due to corrosion, theinduced emf will also decrease according to Eq. 1. The secondary coil located in the center of thesensor measures this induced emf, so that it is possible to determine whether or not section loss
has occurred. The appearance of the developed sensor is shown in Fig. 1.

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Fig. 2. Test set up

Fig. 1. Appearance of EM sensor

3. Non-Destructive Test with the EM Sensor
3.1. Details of the test cable specimens and test program
In order to evaluate the detection performance of the sensor, cable specimens with artificiallydamaged cross-sections were fabricated. The damage area ratio and damage length were used as
test variables. The damage area ratio was set to 1.6, 3.2, 4.8, 6.5, and 8.1 %, and the damage
length was determined to be 20, 50, 80, 120, and 150 mm. A total of 18 cable specimens (4.5 m
length) were produced, including non-damaged specimens. In order to increase the effectivenessof the test, the specimens were fabricated with the identical material as the cable of a cablestayed bridge completed in 2012.
To simulate the environment similar to the actual bridge condition, the NDT was
conducted with the cable specimen installed at an angle in Fig. 2. The sensor was controlled via a
laptop connected wirelessly via Wi-Fi. It was confirmed that it works even at a distance ofseveral tens of meters. The alternating current value is applied to the sensor to generate themagnetic field and the emf is measured by the sensor at a sampling rate of 4,000 Hz.
3.2 Test results
Fig. 3 shows the peak value of the induced voltage along the length direction of the cablespecimen. In the case of the undamaged specimen, the induced emf with a smooth curvature wasmeasured along the longitudinal direction of the specimen. On the other hand, when there isdamage, it could be seen that a rapid decrease in induced emf is measured at the corresponding
location. The tendency of the induced emf according to the degree of cross-sectional loss and thelength of the damage is currently under analysis.

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Fig. 3. Test results of non-damage and three 6.5 % section loss specimen
4. Conclusions
Despite its structural importance, there is no effective non-destructive test method to evaluate thecable's internal condition. Therefore, an EM sensor capable of detecting damage to a cablesection was developed using the law of electromagnetic induction, and its performance was
evaluated through an indoor NDT. As a result of the test, it was found that when the loss occurs
to the cable cross-section, the induced electromotive force generated at the corresponding
location is decreased compared to the undamaged cross-section. If the advanced sensor underdevelopment is achieved, such as applicability verification at the actual sea-bridge site, it isexpected to be of great help in the maintenance of cable-stayed bridges in Korea.
5. References
Jordan, E. and Balmain, K.G. 1968. Electromagnetic Waves and Radiating Systems (2nd edition).
Prentice-Hall
Kwahk, I., Park, K.Y., Choi, J.Y., Kwon, H., and Joh, C. 2020, Non-destructive for sectional lossof external tendon of prestressed concrete structures using total flux leakage, Applied Sciences,
MDPI, 10, 7398.
Seoul Metropolitan Government. 2017. Report on PSC Box Girder Bridges’ Tension Member
Maintenance Plan in Seoul.
Ulaby, F.T. and Ravaioli U. 2016. Fundamentals of Applied Electromagnetics (7th edition).
Pearson Education Korea.
Zhao, J.J. and Tonias, D.E. 2013. Bridge Engineering (3rd edition). McGrawHill Education(Asia).

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Bridge Engineering Institute Conference 2023 (BEI-2023)
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Evaluation of Salt Resistance of Three-Component Cement Concrete by
Unsteady Electrophoresis Test
Kohei Kanaoka1 and Tatsuya Nukushina2*
1: Tokuyama college of technology, Yamaguchi and Japan; email: c17kanaoka@tokuyamakosen.ac.jp
2: Tokuyama college of technology, Yamaguchi and Japan; email: nukushina@tokuyamakosen.ac.jp*: corresponding author

Keywords: chloride attack; Blast furnace cement; Three-Component Cement Concrete; UnsteadyElectrophoresis Test; Effective diffusion coefficient
Abstract: In Japan, marine concrete structures use blastfurnace slag cement type B, which usuallyuses better chloride attack than oridinary portland cement. Although, in the long-term chloride
ingress, control of maintenance costs are increasing due to repair and re -deterioration after repair.
Three-Component Cement Concrete, which replaces the fly ash in the blastfurnace slag cement
type B is considered to have higher salt -shielding proprety than blast furnace cement B concreteand ordinary concrete, but the durability of Three-Component Cement Concrete and thereplacement rate of fly ash,which most salt -shielding propertr are not clear. Therefore, thereplacement rate of fly ash which the most durable of the Three-Component Cement Concrete wasinvestigated using an unsteady electrophoresis test with a short test period and a compression
strength test.
1. Introduction
In order to improve the durability of concrete structures, the use of mixed cements that replace theindustrial by-product and waste in cement is increasing. Above all, ground glanulated blastfurnace
slags (hereinafter, GGBS) are generated as by -products when producing pig iron in the shaft
furnace, long-term strength and salt damage resistance can be expected to be improved.
However, the durability has been lowered due to chloride attack even with the blastfurnace slag
cement type B, which replaces 40% to 60% GGBS from the oridinary portland cement (hereinafter,
OPC). As such control of maintenance costs are increasing due to repair and re -deterioration after
repair. Therefore, the practical application of the Three-Component Cement Concrete which
replaces the fly ash (hereinafter, FA) in the blastfurnace slag cement type B (hereinafter, BB) has
been replaced, has been attracting attention. FA is an industrial by -product that occurs when coalis burned at a coal power plant, and can be expected to improve long -term strength and chloride
attack resistance as well as GGBS. The Three-Component Cement Concrete has not cleared the
durability and the replacement rate of FA, which the most salt-shlieding property．Therefore, inthis study, the purpose was to evaluate durability due to changes in FA replacement

119Bridge Engineering Institute Conference 2023 (BEI-2023)
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Table 1. Formulation of Mortar and experimental conditions
Unit Amount(kg/m3)
Applied Energizeation
NumberW/B
Type
Voltage
time
of
(%)
W
BB
FA
S
(V)
(h)
specimens
BB
365
730
0
BB-FA8361
664
58
BB-FA10360
648
72
BB-FA15
50
358
609
107
1000
30
1
24・48・72
BB-FA20355
568
142
BB-FA30
350
490
210
BB-FA50340
340
340
BB: Blast furnace cement B type (Density:3.04g/m3, Specific surface:3960cm2/g), FA: Flyash (Density:2.23g/m³, Specific surface:3170cm2/g), S: Crushed sand (Density 2.67g/m3)
rates, using the unsteady electrophoresis test that can obtain a concrete chloride penetration
characteristics in a short time indoors and compression strength test.
2. Unsteady Electrophoresis Test
2.1. Materials and Formulation
Table-1 shows the materials and mix proportion of specimens, and the experimental conditions.
Mortar, which is easy to produce a sample under test, is used because the replacement rate of FA,
the most salt-shielding property is not clear. In order to evaluate the salt -shielding properties by
the replacement rate of the FA, the unit amount of the fine aggregate was fixed to 1000 and thewater cement ratio was unified to 50 %. FA was replaced by 8 %, 10 %, 15 %, 20 %, 30 %, and
50 % from the mass of cement. In addition, the name of the subsequent experimental case shall beabbreviated as a combination of the replacement rate of BB and FA, and for example, theFA10%replacement shall be "BB-FA10".
2.2. Test method
Three test specimens of Φ100 × 50 mm dimensions were cut from the center of a Φ100 × 200 mm
test specimen cast by the formulation of Table-1 and used as test specimens. The applied voltage
was constant at 30V. For measurement after the end of the energization, referring to the preceding
research1), after fracting the test specimen, spray a silver aqueous solution (0.1N) on the fracturesurface, and 7points that area colored white were measured in depth, except 10 mm of both ends
of test specimens. The average value was chloride ion penetration depth. Fig.1 shows theexperimental setup of the unsteady electrophoresis test.
The Public Works Research Institute method was used for calculation methods for the execution
diffusion coefficient2). The formula of the Public Works Research Institute method is shown in (1).
𝐷𝐷𝑛𝑛𝑛𝑛𝑛𝑛𝑛𝑛 = 𝑘𝑘 ∙𝑅𝑅(273+𝑡𝑡)
𝑧𝑧𝑧𝑧

𝐿𝐿

(1)

∙ ∆∅

Dnssm : effective diffusion coefficient(×10-12m ² /s) ， t:temperature (℃) ， L: thickness of
specimen (mm)，Xd : chloride penetration depth(mm),R : Gas constant (=8,134J/K・mol)，z:
Absolute value of chloride ions (=1)，F: Faraday constant (=9.648×104J/V・mol)，∆∅: applied

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Chloride penetration depth(mm)voltage (V), The penetration speed K was calculated based on the inclination of the approximation
curve that passed through the origin in the relationship between the applied time and the chloride
penetration depth.
50
40
30
20
OPC3)
BB-FA8BB-FA15
BB-FA30

10
0

0

24

48

BB
BB-FA10BB-FA20
BB-FA50

72

96

energizeation time(h)Fig. 1. Unsteady electrophoresis test

Fig. 2. Chloride penetration depth

2.3. Test results and consideration
Fig. 1 shows the relationship between the applied time and the chloride penetration depth. The
case of OPC is referred to the previous study 3). In any mixture, the longer the applied time, themore chloride permeation depth. Also, the larger the replacement rate of the FA, the more thechloride penetration depth. The penetration depth at the same applied time is BB-FA10, BB-FA8,
BB, BB-FA15, BB-FA20, BB-FA30, BB-FA50, OPC were increased in the order. chloride
permeation depth of the BB-FA50 was measured a test body that the was energized 24,48 hours,
because the entire fracture surface was colored white when the sliver nitrate solution was sprayedon the fracture surface at the end of the power to 72 hours.
Fig 2 shows the calculation result of the effective diffusion coefficient. Fig 3 shows the3

4.5

Effective difficient coefficient(cm²/year)

Effective diffusion coefficient(cm²/year)

5.0
4.0
3.5
3.0
2.5
2.0
1.5
1.0
0.5
0.0

OPC3)

BB

BBFA8

BBFA10BBFA15

BBFA20BBFA30

2.5
2
1.5
1
0.5
0

BBFA50Fig. 3. Effective diffusion coefficient

0

10

20

30

40

Replacement rate of FA(%)

50

Fig. 4. Effective diffusion coefficient andthe replacement rate of FA

relationship between the effective diffusion
coefficient and the replacement rate of FA.

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These figures indicate that BB-FA10 has the lowest effective diffusion coefficient. Also, the larger
the FA replacement rate, the larger the effective diffusion coefficient.
From these results, it can be seen that replacement rate of 10% by FA has most saltshlieding property. And, the larger the replacement rate of the FA, the lower the amount of calcium
hydroxide required for the reaction. Decreasing calcium hydroxide means the amount of unreacted
BFS and FA are increasing and that the densification of void may have stagnated.

40
35

40

32.3
28.4

30

29.7

Compression strength(N/mm²)

Compression strength（N/mm²)

3. Compression Strength Test

30.9

30

25.0

25

21.2

20

20

13.6

15

10

10
5
0

BB

BBFA8

BBFA10BBFA15

BBFA20BBFA30BBFA50

0

0

10

20

30

40

Replacement rate of FA(%)

50

Fig. 6. Replacement rate of FA andcompression strength

Fig. 5. Compression strength test

3.1. Formulation and test method
The test specimen used the formulation of Table-1. Three specimens were produced withdimensions of φ50 x 100mm for each formulation. The test specimen was demolded at 7 days ofdimensions of φ50 x 100mm for each formulation. The test specimen was demolded at 7 days ofthe material age, and performed underwater curing until 28 days of the material age, and thenperformed a compression strength test. It was performed in accordance with JIS A 1108(JapaneseIndustrial Standards)．
3.2. Test Results and Consideration
Fig.5 shows the compression strength test result. Fig. 6 shows the relationship between FA
replacement rate and compression strength. These figures indicate that the compression strengthtended to decrease as the replacement rate of FA increased. On the other hand, the decrease instrength was minor up to about 15 % FA replacement. From this, it is considered that the higher
the FA replacement rate, the lower the cement amount, and the strength did not appear.
4. Conclusions
In this study, the following findings were obtained by quantitatively evaluating concrete durability
through the unsteady electrophoresis test and compression strength test.
(1) The most highly salt -shielding property is that 40% of the blast furnace slag powder is replacedwith the OPC and the FA is replaced by 10%.
(2) The larger the replacement rate of FA, the greater the chloride penetration depth and theeffective diffusion coefficient.
(3) The larger the replacement rate of the FA, the lower the compression strength.

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5. Reference[1] Takeshi IYODA et al, “Evaluation for curing effect on blast furnace slag concrete by unsteadystate electrophoresis test”, Cement Science and Concrete Technology, Vol.68, pp.275-282,2014[2] Yutaka WATANABE et al, “Chloride ion diffusion coefficient of concrete by rapid chloride
penetration test calculation of the coefficient of chloride diffusion”, Concrete Engineering Annual
Proceedings in Japan, Vol.24, No.1, pp.663-668,2002[3] Ryota KUMAMOTO et al, “Evaluation of Salt Resistance of Highly Durable Concreteby Unsteady Electrophoresis Test”, Japan Society of Civil Engineers,2022

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Field TestingBridge Engineering Institute Conference 2023 (BEI-2023)
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Grating on Bridge deck for Improving Flutter Performance: A Case Study ofCable-Suspended Bridge
Zhen Wang1*, Xuhui He2, Haiquan Jing3, and Lulu Liu4
1

: Department of Civil Engineering, Central South University, Changsha and China; email:wangzhen@csu.edu.cn
2: Department of Civil Engineering, Central South University, Changsha and China; email:xuhuihe@csu.edu.cn
3: Department of Civil Engineering, Central South University, Changsha and China; email:hq.jing@csu.edu.cn
4: Department of Civil Engineering, Central South University, Changsha and China; email:liululu2020@csu.edu.cn*: corresponding author
Keywords: cable-suspended bridge; flutter; aerodynamic countermeasure; wind tunnel test;
grating
Abstract: Wind tunnel tests were conducted to investigate the flutter performance andcharacteristics of a cable-suspended bridge girder at various angles of attack (AOAs). The results
showed that the prototype bridge deck had poor flutter performance and was highly sensitive tovariations in AOAs. For improving the flutter performance of the girder, the aerodynamic
countermeasure of grating-installed on bridge decks with various grating-opening ratios (gratingORs) was also conducted in wind tunnel tests. It was found that the grating-installed bridge deck
was an effective way to increase the flutter critical wind speed (FCWS) of a cable-suspended
bridge. The energy participation ratio (EPR) was considerably changed by the grating installedon the bridge deck. The EPR of both vertical and torsional motions shows high correlations with
FCWSs, which may explain the improvement of FCWS.
1. Introduction
The cable-suspended bridge is a flexible system with cables as the main load-bearing structure
and the deck laid directly on main cables, and the deck alignment is basically the same as themain cable alignment, as shown in Fig. 1. Its strong spanning capacity, fast erection speed, lowengineering investment and construction difficulty, especially suitable for geological conditions
in mountainous areas. Therefore, cable-suspended bridges play an increasingly important role inaddressing the transport problems of mountain residents. Despite having many benefits, cablesuspended bridges perform poorly in wind resistance. Moreover, the wind environment thatbridges must face with in mountainous regions is complex, with characteristics like large AOAs,
non-uniform, and non-stationary. As a result, the poor wind resistance performance of cablesuspended bridges is becoming more and more evident and is a significant technical barrier totheir promotion and application. Several cable-suspended bridges in China have reportedly been
subjected to wind-induced vibration in recent years.

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In the present study, a cable-suspended bridge, which has been subjected to flutter on 2021, wasserved as the research object. Since aerodynamic countermeasure is an economical and effective
way to improve the aerodynamic stability of bridges (Ge and Xiang, 2009; Sato et al, 2002), an
aerodynamic countermeasure of grating-installed on the bridge deck was selected. Experimental
approaches were primarily considered when determining the flutter performance of the cablesuspended bridge. Flutter characteristics, including FCWSs and EPRs were also investigated.

Fig. 1. Cable-suspended bridge.
2. Wind Tunnel TestWind tunnel tests were carried in the wind tunnel of Central south university, and theinformation of the wind tunnel was reported by (He et al, 2018). The geometric scaling ratio of
the section model is 1:10, the length is 1 m and the wind speed ratio is 2.08. The specific design
parameters of the section model are shown in Table 1, and bridge decks with grating-OR of 040% at the AOA of ±5° were tested. In wind tunnel tests, the wind speed was measured using a
Cobra probe. The displacement of the section model was measured by displacement signals
using four Keyence laser transducers with a measuring range of 30 cm and a sampling frequencyof 1000 Hz.
Table 1. Physical properties of section model
NO.
1
2
3
4
5
6
7
8

Items
Mass
Inertia
Vertical frequency
Torsional frequency
Frequency ratio
Vertical damping ratio
Torsional damping ratio
Wind speed ratio

Unit
kg/m

kg·m2/m
Hz
Hz
/
%
%
/

Prototype1404.290
10922.850
0.360
0.429
1.193
/
/
/

Scale ratio
1/10
1/10000
5
5
1
1
1
1/2

Target value
14.043
1.092
1.798
2.145
1.193
0.350
0.350
/

Actual value
14.165
1.107
1.734
2.166
1.249
0.380
0.380
/

3. Response Characteristics
3.1. FCWS
Fig. 2 illustrates the FCWSs of wind tunnel tests for the grating-installed bridge deck. The
FCWS considerably increased when the grating-OR increased. Particularly when the grating-OR
increased to 30%, the FCWSs at various AOAs meet the requirements of flutter checking windspeeds. As the grating-OR reached 30 %, not flutter occurred at the highest test wind speed (70m/s) for negative AOAs. The range of AOA without flutter at the maximum test wind speed was
extended to 0° as the grating-OR reached 40%. For positive AOAs, the FCWS increases with126

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increasing grating-OR up to 20%, but then remains almost constant for higher grating-ORs. It isconceivable that there is an optimum grating-OR and that once achieved, the FCWS essentiallyremains unchanged.

Wind speed (m/s)

80
60
40
20
0

0%
10%
20%

-6

-4

30%
40%
Flutter checking wind speed-20
2
AOA (°)

4

6

Fig. 2. Test results of FCWSs
3.2. Energy participation ratios
To further analyze the participation level of vertical and torsional degrees of freedom during theoccurrence of flutter, the EPR was introduced to evaluate the flutter pattern. The EPR essentiallyrepresents the distribution of vibration energy in the vertical and torsional degrees of freedom
and reflects the ratio of mechanical energy in the vertical and torsional motion of the total systemmechanical energy, which can be calculated by the following equation:(1)
where, Ψ is the average amplitude ratio of vertical motion to torsional motion; η is the ratio of
the radius of gyration r to the half-width b of the bridge deck.
Fig. 3 shows the variation of the EPR with the grating-OR at various AOAs. It can be seen that
the EPR was influenced by both AOAs and grating-ORs. For the prototype bridge deck, Ph isgreater than Pα at negative AOAs, the reverse is true at the AOA of 0°, and the latter isdominant at positive AOAs. The FCWS increases as Ph increases, and gets the lowest value at
positive AOAs where Ph is at the lowest levels. For the grating-installed bridge decks, Ph
increases with the increase of grating-ORs in most cases, and the same with the FCWS. Similar
results were obtained by (Yang et al, 2015) in their study of the aerodynamic stability a twin box
girder with various slot widths, i.e., the participation level of the vertical motion was highly
correlated with the FCWS. It indicates that the EPR is one of important factors determining the
FCWS. On the other hand, Ph increases rapidly with grating-ORs at negative AOAs and 0°
AOA, while that increases slowly at positive AOAs. Ph attains a high level for a grating-OR of
10% at negative and 0° AOAs, whereas it attains an equivalent level for a grating-OR of 30% or
40% at positive AOAs. This may explain the reason why the FCWSs of negative AOAs and 0°
AOA exceed the highest wind test speed with increasing of grating-OR , while the FCWSs of
positive AOAs remain lower than that.

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4. Conclusion
In the present study, for improving the aerodynamic stability of a cable-suspended bridge, aseries of wind tunnel tests of the bridge deck with various grating-ORs and AOAs was conducted.
The main conclusions can be drawn as follows:
1 The aerodynamic countermeasure of grating installed on the bridge deck significantly
improved the flutter performance of the cable-suspended bridge. The FCWS increased as thegrating-OR increased, but there was an optimal grating-OR. After reaching the optimal gratingOR, the FCWS increased very gently, especially for positive AOAs.
2 The EPR was effectively influenced by the grating-installed bridge deck at the same AOA. As
the grating-OR increased, Ph significantly increased while Pα dramatically decreased, which was
one of the key factors in enhancing flutter performance.

10
20
30
Opening rate (%)

10
20
30
Opening rate (%)

(b)

Ph
Pα
FCWS

20
10
20
30
Opening rate (%)

400Wind speed (m/s)

40

0.4

Ph
Pα
20
FCWS

0.2
0.0

0

10
20
30
Opening rate (%)

(c)

Ph
Pα
FCWS

0.8

40

0.2

20

1.0

60

0.4

0

40

80

0.6

0.0

0

(a)

0.8

EPR

0.0

0

EPR

1.0

40

Ph
Pα
FCWS

0.2

Wind speed (m/s)

0

20

40

0.4

60

0.6

80

40

0.4

0.0

0

60

0.6

20

0.2

0

40

Wind speed (m/s)

Ph
Pα
FCWS

0.2

0.6

80

0.8

60

EPR

40

0.4

EPR

0.6

1.0

80

0.8

60

Wind speed (m/s)

EPR

0.8

0.0

1.0

80

Wind speed (m/s)

1.0

0

10
20
30
Opening rate (%)

40

0

(e)

(d)

Fig. 3. EPRs at various AOAs: (a)-5°; (b)-3°; (c)0°; (d)3°; (e)5°5. References
Ge, Y.J., Xiang, H.F., 2009. Aerodynamic stabilization for box-girder suspension bridges withsuper-long span, in: Proceedings of the In Proceedings of the 5th European and African
Conference on Wind Engineering. Firenze University Press.
Sato, H., Hirahara, N., Fumoto, K., et al., 2002. Full aeroelastic model test of a super long-spanbridge with slotted box girder. Journal of Wind Engineering and Industrial Aerodynamics 90,
2023-2032.
Yang, Y., Wu, T., Ge, Y., et al., 2015. Aerodynamic Stabilization Mechanism of a Twin Box
Girder with Various Slot Widths. Journal of Bridge Engineering 20, 04014067.

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Development of Elevated Road and Rail Projects in Singapore
Hooi Leng Phua1*
1: Land Transport Authority, Singapore; email: phua_hooi_leng@lta.gov.sg*: corresponding author

Keywords: Jurong Region Line; Integrated Hub; Noise Barriers; Viaduct Segments
Abstract: Since the opening of Singapore’s first Mass Rapid Transit (MRT) Line in 1992, thesystem has grown from the initial 67 kilometers and 42 stations to 240 kilometers and 175 stations
presently, and will expand by 50% to 360 kilometers by 2040. Like many urban cities, Singapore
is a built-up city state with very limited land space to meet its varied needs. Majority of the MRTlines and road expressways that were implemented in the past were constructed underground to
free up the above-ground spaces for other developments. Whilst this intensifies the development
potential and land value capture, the construction and operating costs for building underground isvery prohibitive, and not necessarily always justifiable. To ameliorate the footprint andenvironmental impact of above-ground transport infrastructure on the landscape, deliberate effort
is made at the planning and design stage to integrate them into developments to intensify theutilisation of the land.
1. Introduction
With a land area of only 730 square kilometers, Singapore is one of the densest cities in the world
with a population density of over 8,000 people per square kilometer. A fine balance has to be
struck amongst all the different competing demands for limited land, which is atypical of many
developed cities. 12% of Singapore’s land is already used for roads, almost the same used for
housing. It is not sustainable for Singapore to continue to expand its road network to support thegrowth and development of its urban spaces. Singapore’s land transport policy is centered onproviding an efficient, safe, affordable public transport system that meets the needs ofSingaporeans. With a robust and comprehensive rail network as the backbone, Singapore
envisages to grow the existing rail network to 360km by 2030, such that 80% of all households
can be within 10-minutes walking distance from a MRT station.
However, the construction of a transport system is a high cost investment that can take 7
to 10 years to construct. The cost and duration required to implement it is further compounded if
it is underground. Hence, utiltising underground space for transport infrastructure is very common
in highly urbanised cities to minimise land sterilisation and maximise the development potential
and value of the land above, is a careful balance planners must achieve.
2. The Jurong Region Line
Singapore is currently constructing the seventh line - Jurong Region Line (JRL) to provide directconnectivity between the North-Western and Western part of the island upon phase completionfrom 2027 (see Fig. 1). Comprising of 24 stations, an integrated depot and 24 kilometers route
length, it is a fully above-ground medium rail system that serve existing and upcoming

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developments under the overall longer-term plans to transform Singapore’s western business
district.

Fig. 1. Jurong Region Line Railway Map
Being elevated allows the stations to be closer-spaced and improve the accessibility of key
activities nodes in the western region, serving commuters better. Without the need for extensive
deep excavations next to adjacent buildings, sensitive structures and under utilities, it allows theconstruction period and cost to be greatly reduced.
However, there are numerous engineering challenges that have to overcome. Some of thestructures of existing operational Boon Lay, Choa Chu Kang and Jurong East stations, with highdaily ridership, have to be demolished and strengthened. Sections of the JRL viaducts will be
constructed over busy roads, existing rail lines, and close to existing buildings. In addition, global
supply chain disruptions, shortage of skilled manpower and material shortages have added otherchallenges to the construction works.
To help overcome these, digital twins are created to allow the architects and engineers tovisualise and optimise the layout of the stations and commuter flow. Critical construction activities
are modelled virtually, with each step of the works simulated to allow for a better understandingof the work sequence, risk identification and mitigation. Pre-fabrication and pre-casting are also
used to form the station and viaduct elements. The viaducts comprise of about 12,000 precast
segments, each weighing 20 to 40 tonnes, which are fabricated off-site, before transporting themto the required location to be lifted and placed by lifting girders (see Fig. 2(a)) or launching frames(see Fig. 2(b)). These methods allow the viaducts to be constructed faster, with lesser manpower
and also reduces the disruptions and inconveniences to the communities where the works are being
carried out in.

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Fig. 2(b). Balanced Cantilever using
Lifting Frames

Fig. 2(a). Segmental Launching using
Lifting Girder

Conscious effort was made at the onset of its planning to integrate with existing and new
developments to maximise the development potential of the corridor. The existing Jurong East
Station, which already serves as an interchange of two existing lines will be expanded with theaddition of a pair of stacked platforms to allow seamless transfer with the JRL. Instead of having
two separate stations with more land take, the single expanded station with a common transfer
level will connect all the six platforms without requiring commuters to exit and re-enter the
faregates as they interchange between the three lines.
One environmental downside of an elevated railway line is the constant and permanentnoise emitted from the running rails, which can be very debilitating to residents living next to thetracks. To overcome this, permanent noise barriers (see Figs. 3(a) and (b)) will be installed along
the viaducts that are near residential areas.

Fig. 3(a). Typical 2m height Noise BarriersFig. 3(b). Typical 5m Height Portal
Type Noise Barriers at Track Turnout

3. Integrated Hub at Jurong East
The land parcel on the Northern end of Jurong East station, where the viaducts leading to theexisting station crisscross, is being concurrently developed with a new 27-storey, 2-level basement,
mixed use development comprising Grade-A office space, retail and civic facilities and a bus
interchange that will integrated with the expanded station. When completed by 2028, it will forma single Integrated Hub (see Fig. 4) in the center of the new Jurong Lake District. By constructing
this new development together with the expansion and alteration works to the station, it minimises
the overall duration and disruption to the operations of the station, and also obviates the added
constraints that will be poised by the operations of a new completed line if constructedindependently. Underpinning every aspect of the design and construction of this new development

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is needed to maintain the operations and safety of the station and the five railway viaducts that are
cutting through the plot.
When completed, a new four-storey Hub Plaza (see Fig. 5) will be the centroid of theexisting station, new buildings and existing adjacent developments. This Hub serves to connect all
the developments and station platforms both multi-directional, and also at different levels.
Positioning the Hub Plaza away from the existing platforms was a key design consideration as the
new transfer platform that is within it would not need to be constructed above the existing
operational platforms, which would cause disruptions to the daily commuters and result in highrisks to the operations of the station.

Fig. 4. The new Integrated Hub

Fig.5. Sectional Layout of the Hub Plaza

4. North-South Corridor
The North-South Corridor (NSC) was originally conceptualized as a vehicular expressway toconnect Northern Singapore and the city centre directly. The 21.5 kilometers expressway will
improve the transport connectivity of the towns along this corridor. NSC is also Singapore’s
longest Transit Priority Corridor, featuring dedicated bus lanes, cycling trunk routes and pedestrianpaths to serve public bus commuters, active mobility users and pedestrians. This is a shift awayfrom typical expressways in Singapore, which are dedicated to moving large volumes of private
motor vehicles around the island, to embracing multi-modes of transport within a shared corridor.
Whilst majority of the NSC is underground as it meanders through the highly-densified corridor
in the South, there is 8.8 kilometers of elevated road viaducts in the North where it is less developed.
5. Conclusion
Transport has always been one of the most important pillars of any modern economies, albeit also
one of the most challenging. Planners must grapple with increasing demands for efficiency,
security, reliability, cost-effectiveness and lately also care for the environment. These demands are
made more challenging with the marked increase in projected ridership arising from population
growth and rising affluence. The Land Transport Authority of Singapore has to continuously
orchestrate long term integrated planning and optimise the infrastructure to sustain the continued
development of Singapore’s Road and Rail system.
6. References
Land Transport Authority 2013. Land Transport Masterplan 2013, 20-21

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Structural Behavior of a 40+ Year Old PSC-I Concrete Bridge
According to its Internal Tendon Loss in the Central Part
Yeonghun Seong1 and Hyunjoong Kim2*
1: Hybrid Structural Testing Center, MyongJi University, Yongin, Republic of Korea; email:syh0896@mju.ac.kr
2: Hybrid Structural Testing Center, MyongJi University, Yongin, Republic of Korea; email:khjmju@mju.ac.kr*: corresponding author

Keywords: PSC-I, Tendon, Rupture, Full-scale Experimental, Maintenance
Abstract : Prestressed concrete bridges with a type I girder (PSC-I), hastily built in the 1960s,
have been increasingly deteriorating, making their internal tendon prone to rupture, leading tosocial and economic problems due to bridge collapse and emergency repairs. In this study, apost-tension PSC-I bridge, to be demolished after more than 40 years of service, was evaluatedfor its structural behavior due to internal tendon loss. The internal tendon of its central part, aweak flexural behavior part, was artificially cut to model an actual damaged structure; its
structural behavior was then compared to that of an intact girder. The results showed thatstiffness increased from 2.2% to 37.9% depending on the amount of internal tendon loss (lossratio from 8.3–25%) in the central part of the bridge, with a load reduction of up to 78% whencompared to an intact girder. The position of the neutral axis and maximum progress of the crack
also increased rapidly due to the change in stiffness and decrease in internal force.
1. Introduction
In the 1960s, as Korea experienced rapid social development, new bridges were being
constructed as fast as possible. However, these bridges now face maintenance problems.
Prestressed concrete (PSC) bridges with internal tendons account for 28% of all domestic bridges.
Due to their structure, internal tendons are causing particularly pressing issues in the social andeconomic areas, including rupture or section reduction from corrosion and deterioration. Tests on
the structural behavior of PSC bridges before and after the rupture of their internal tendon havebeen conducted at home and abroad. However, upon examining previous research, it was foundthat it was limited to evaluating the design values and simple beam behaviors of aging bridges
that have been in existence for over 40 years. In addition, when verifying the effect of internaltendon fracture, a new girder was created and performance evaluation was conducted based on
the internal tendon fracture effect and loss amount by location. Therefore, research on structural
performance evaluation is needed for aging bridges that have been in existence for over 40 years,
specifically focusing on the loss amount of internal tendon by location in the central part of thebridge.

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2. Experimental Method and Contents
In this study, a structural test of a over 40 years old PSC-I girder was concducted to understandthe bridge’s structural behavior based on the loss of the internal tendon of the girder during
actual use, the central internal tendon of a bridge to be tested was artificially ruptured to allow
monitoring of the structural behavior changes according to the internal tendon loss rate(from 0 to25%) and comparing them with an intact structure. Test model specifications and details of theexperimental load cases and loading process are in Fig 1. Fig 2. and Table 1.

Fig. 1. Cross section of a PSC-I girder bridge and Sensor set-up layout*Loss Ratio = Ruptured Steel Wire / 48 (=4 Tendons x 12 Wires)Fig. 2. Schematic of the artificial rupture of the internal tendon
Table 1. Details of the experimental load cases and loading processP
Loading
Load Case and parameters tested
Remarks(kN)
Process
Elastic behavior
150
3 times
60% of computation crack loadReference Girder
Crack load300
1 time120% of computation crack load
(Ref. Girder)
Pmax behavior
869
1 timeElastic behavior
150
3 times
60% of computation crack load
Crack load300
1 time120% of computation crack load
Cutting Girder
Nonlinear behavior
600
1 time(8.3% Loss andCutting strand
600
1 time*Strand loss ratio: 8.3 %25% Loss)
Cutting strand
600
1 time*Strand loss ratio: 25 %Pmax behavior
679
1 time-134

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3. Evaluation of the Flexural Behavior of A PSC-I Girder
Girders prepared for this study were tested for their maximum load, maximum displacement,
concrete strain, and steel wire strain, as shown in Table 3, for three different loss conditions toevaluate the flexural behavior of a girder in use for more than 40 years and identify the
behavioral characteristics in case of internal tendon rupture. The flexural behavior of the girdervaried according to its internal tendon loss rate, as observed by the differences in stiffness after
the crack load, which varied between 2.2–38%. However, because the experiment of the girderwith an 8.3% loss ratio ended at the target load of 600 kN, the deviation of stiffness wascompared and evaluated in the same load range.
Table 2. Test results
Experimental
P (kN)
condition
Ref. Girder
848.95Girder with
600.0
8.3% Loss
Girder with
680.0
25% Loss

Disp. (mm)εcεtEI (Stiffness)

181.6

1,496

10,403

1

66.0

735

2,853

0.97

200.3

2,835

8,146

0.62

(a)

(b)

(c)

(d)

Fig. 3. Behavior of 45 years PSC-I girder : (a) Load–Displacement Relationship; (b) Load-toConcrete Stress Relationship (Compression Part); (c) Load-to-PS Steel Wire Stress Relationship(Tension Part); (d) Neutral axis position and crack height change

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4. Conclusion
In this study, the load-carrying capacity for the service load and ultimate load of an old PSC-I
bridge in operation for over 40 years was examined. The flexural behavior characteristics,
according to the loss rate of the internal tendon of the bridge, were identified by artificiallyrupturing its internal tendon. The following conclusions could be drawn by analyzing the change
in stiffness through maximum load displacement, the change in material properties for each loadstage through the change in strain in the compression and tension parts, and the crack progress
pattern according to the change in the position of the neutral axis.
1) Old PSC-I bridges, as all bridges in general, show rapid displacement changes while
experiencing crack loading. In this study, after the steel wire yielded, depending on theloss rate of the steel wire in the tension section (0 to 25%), a nonlinear displacement of
up to 37.9% was observed.
2) The concrete in the compression part and steel wire in the tension part both generated astrain difference from 27.7% to 93.5%, depending on the loss rate of the steel wire. The
plastic state that occurred after the steel wire yielded caused a rapid increase in thestrain experienced by the concrete.
3) The flexural crack advanced up to the vicinity of the upper flange when the crack load
was exceeded, and the position of the neutral axis and the height of the crack developed
as the loss rate of the internal tendon was increased.
4) When the internal tendon is ruptured by corrosion—a frequent occurrence in many old
PSC bridges—and a load greater than the service load is applied, a rapid change instructural behavior occurs depending on the internal tendon loss rate.
This study examined the flexural behavior characteristics of PSC-I bridges according to theamount of loss of the internal tendon in the central part. Further work is necessary to evaluate theflexural behavior according to internal tendon rupture for each PSC-I girder position. This
information must be used to perform maintenance work on mid- to long-term bridges by
predicting their behavior by FEM analysis, a reliable method to acquire information on themechanics and durability characteristics of its constituent materials along the service life ofbridges.
5. Acknowledgements
This work was supported by the Korea Agency for Infrastructure Technology Advancement(KAIA) through Grant 21CFRP-C163381-01, funded by the Ministry of Land, Infrastructure andTransport of Korea.
6. References[1] AASHTO. 2014. Bridge design specifications. Washington, DC: AASHTO.
[2] ACI Committee 228 (2004), Nondestructive test methods for evaluation of concrete instructures, Report ACI 228.2R-98, American Concrete Institute, Farmington Hills, MI.

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[3] Pessiki S., Kaczinski M., Wescott H. Evaluation of Effective Pre-stressed Force in 28-YearOld Prestressed Concrete Bridge Beams. PCI Journal. 1996; 41(6): 78-79.
[4] Shenoy C., and Frantz C. Structural Tests of 27-Year-Old Prestressed Bridge Beam. PCI
Journal. 1991; 36(5): 80-90.

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The Use of Engineered Cementitious Composite (ECC) Deck Overlay at
Curry Street Bridge, Windsor, Ontario, Canada: Lessons Learned andChallenges
Don Gardonio1, Steve Volpatti1, Emad Booya2, and Philip Loh3,41Facca Incorporated, Ruscom, Ontario, Canada
University of Windsor, Windsor, Ontario, Canada
3
Dura Concrete Canada Incorporated, Ruscom, Ontario, Canada
4philip@facca.com
2*: corresponding author

Abstract: Traditional High Performance Concrete Overlays for bridge decks deteriorate due tomany factors. These factors can be due to excessive loading, shrinkage, weathering effect, etc.
The replacement cost of existing deficient concrete pavement is expensive. Moreover, the designof maintenance materials requires the use of energy-efficient materials with a low environmental
impact. Facca Incorporated, in collaboration with Dura Concrete Canada Inc., a change proposal
was approved by the City of Windsor to replace the high performance concrete overlay by a highperformance fibre reinforced composite (HPFRC) known as Engineered Cementitious
Composite (ECC) material for use as an overlay, as part of the deck rehabilitation for Curry
Avenue Bridge over the Grand Marais Drain in Windsor, Ontario. Using the Dura® ECC overlay
developed by Dura Concrete Canada Inc. and Facca Inc., this innovative, bendable, and durable
ECC product was installed successfully. The visual observations and inspections after more than
8 months in service revealed that there were no signs or indications of deterioration to therepaired Bridge overlay. This paper will provide details of the innovative design, field testing,batching process, the challenges encountered during the placement and the in-service field
conditions after 8 months.
1. Project Background
The City of Windsor tendered a project to rehabilitate the Curry Avenue Structure over theGrand Marais Drain as per the Engineer’s determination. The bridge is a simple span (16m)
consisting of prestressed box beam girders with a concrete topping and a ductile highperformance concrete. The deck cross fall is 2% across the length of the structure. Since thetopping is to be an exposed deck, the new overlay was to be high performance concrete. The
deck was to be scarified 40 +/- 5 mm and then the topping removed locally at spalled anddelaminated areas. The new high-performance overlay material in sections where the topping
was removed was calculated to be up to 100mm in depth. This topping was to be reinforced with
10M dowels at 2.0m intervals and staggered every 1.0m and one layer of welded wire mesh was
to be placed over the entire surface. Facca Incorporated was retained as the general contractor for
the City of Windsor.

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2. Innovation Change Proposal
During the pre-construction meeting, Facca Incorporated requested from the City of Windsor to
replace the traditional High Performance Concrete mix for the deck overlay with by a highperformance fibre reinforced composite (HPFRC) known as Engineered Cementitious
Composite (ECC) material for using as an overlay due to its superior qualities in strength,
durability and performance. Specifically, the mix proposed would have a longer life span, reachearly strengths at a faster rate than a traditional mix and would far surpass the 50 MPa strength
requirements at 28 days set for this contract. The City accepted the proposal with conditions.
3. Dura® ECC Mix Design
3.1.
Mixture Design and Proportioning
High-Performance Fibre Reinforced Concrete Dura® ECC with a minimum 28-day strength of 60
MPa, was designed and implemented as repair materials. The mixture proportions with respect tocement mass are listed in Table 1. The fibres content was 2 percent by volume. As per the roadconditions and specific requirements, the mixtures were designed to have thixotropic consistency.
Thixotropic cementitious mixtures are known to be thick and viscous in normal condition but
flow when subjected to external vibration or agitation.
Table 1. Mixture Proportioning with respect to cement mass
Material ID Cement SCM
Sand
Water
®
Dura ECC 1
1.15
0.89
0.52

HRWR
0.015

WM
0.008

4. Dura® ECC Trial Batch
As part of the requirements of the change proposal, Facca Incorporated was to provide a trial
batch of 2 m3 and cast the mix into a mold of 2.0 m x 4.0 m x 0.2 m depth to evaluate theworkability and finishing properties. This test panel also required a 2% slope to simulate thebridge deck crossfall. Figure 1 shows the trial batch form.

Fig. 1. Form for the Dura® ECC Trial Batch

The trial batch was completed on April 30, 2021 in the presence of the Consultant Engineer andthe City of Windsor Representative. The specified slump flow specified for this mixture was 200
mm to 250mm. Strength cylinders as per CSA A23.1/23.2 were fabricated as required from theplaced trial batch to evaluate the strength. Figures 2 and 3 show the steps for fabricating a
specimen from the trial batch ECC mix. The material thixotropic properties were confirmed andthe material was able to stand up with the 2% slope without slipping or sagging. Curing
measures of wet burlap, opaque poly and insulated tarps were installed.

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Fig. 2. Filling the Trial Batch Form

Fig. 3. Levelling of the Dura® ECC mix on 2% slope (LeftSide is low end)

The compression test results confirmed to the City the rapid and early strength gain and thehigher than required 28-day required strengths. Generally, this trial batch met all performancerequirements as well as the fresh properties (i.e., levelling) requirements, but the finishingrequirements were yet to be confirmed.
5. Dura® ECC Overlay Placement
Placement of the new concrete overlay and patches to the box beam girders took place on May
29, 2021. Ambient temperature at the start of placement was 11°C, the day was sunny andrelative humidity for the day hovered around 40%. After correction of screed elevations with theConsulting Engineer, the total depth ranged from 60mm to 130mm, and a total quantity of 17m3
of ECC were placed. Photos of the surface preparation and placement are shown respectively in
Figures 4 and 5. The concrete was placed by concrete bucket and screeded using an air driven
Allen Screed to level the overlay. Spiked rollers and tool placers were used to maneuver theconcrete ahead of the screed. No concrete vibrators were used due to the plasticity of the mix.
After the screed had passed, the modified tine rake was to be used to groove the surface
perpendicular to the direction of traffic. However, due to the low relative humidity, once thescreed passed the Dura® ECC, the surface was drying out too quickly. Therefore, the decision
was made with the Consulting Engineer and the City to immediate apply the wet burlap and poly
and to grind and groove the surface in the hardened state.

Fig. 4. Curry Avenue Surface Preparation

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Fig. 5. Curry Avenue ECC Placement

Compressive cylinders were taken to track the strength gain throughout the curing and post
curing period. All results for early breaks and 28-day strength (87 MPa average) far surpassed
expectations, reaching design strength at 4 days (56 MPa average). As part of the completion of
the deck overlay, the surface was diamond grinded, and grooves were cut in.
6. In-service Performance Dura® ECC
The visual observations and inspections after more than 8 months in service revealed that there
were no signs or indications of deterioration to the repaired Bridge overlay. The bridge issubjected to heavy traffic volume since placement and exhibited several freeze-thaw cycles anddry-wet cycles. It is also worth noting that the bridge and Dura® ECC material was subjected tosnow removal activities which include salting and/or plowing. Only hairline cracking has been
identified on the surface that are below 0.3 mm in width. These types of cracks are not of aconcern as per CSA A23.1.
7. Conclusions and Lessons Learned
For this project, Dura® ECC has proven to be a viable alternative to traditional HighPerformance Concrete. Factors to consider would include:- Placement of this mix can be achieved using traditional methods, equipment, and machinery.
- Both early strength and target strengths can be achieved, faster and higher than regular highperformance concrete.
- The materials is impacted of not only ambient temperature but also relative humidity during
and after placement.
- Finishing requirements and alternative methods to achieve specified surface finish, wereconfirmed.
- Dura® ECC materials can sustain heavy traffic loading and extreme weathering exposure
despite the age of the patch repair.
8. References
ASTM 2010. ASTM C 618, Standard Specification for Coal Fly Ash and Raw or Calcined
Natural Pozzolan for Use, Annual Book of ASTM Standards.

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ASTM 2013. ASTM C989, Standard Specification for Slag Cement for Use in Concrete andMortars, Annual Book of ASTM Standards.
ASTM 2016. ASTM C1116, Standard Specification for Fiber-Reinforced Concrete, Annual Book
of ASTM Standards.
ASTM 2015. ASTM C494. Standard specification for chemical admixtures for concrete, Annual
Book of ASTM Standards.
Booya E., Adesina A., Gardonio D., Loh P., Das S., Mechanical and Durability Performance ofan Innovative In-Situ Concrete Pavement Repair Patch, Transportation Association Canada,
2020.
CSA 2013. CSA A3001. Cementitious materials used in concrete, Canadian Standards
Association, Mississauga, Ontario
CSA 2019. CSA A23.1/A23.2. Concrete materials and methods of concrete construction/Test
methods and standard practices for concrete, Canadian Standards Association, Mississauga,
Ontario142

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Transporting Heavy Weight Cargo, Diagnosis Philosophy, and a Case Study
Daniel Elmaleh* and Yehonatan Pestes 2
1
2: Via Bridges design and engineering, Tel Aviv Israel; email: Daniel@via-bridges.com: Via Bridges design and engineering, Tel Aviv Israel; email: Yoni@via-bridges.com*: corresponding author

Keywords: Heavy weight cargo; Bridge loads; Bridge level gauges; Live load test
Abstract: Bridges are designed and built to meet specific load criteria. Despite safety margins indesign, an overloaded transport cargo or vehicle may affect the sustainability of the structure andits life expectancy. However, the need to transport special cargo exceeding the design criteria
could arise often. Such heavy-load transportation refers to the movement of vehicles that exceed
the weight limits set by state or federal regulations. This type of transportation is often necessary
to deliver oversized or overweight goods, such as heavy machinery, industrial or construction
equipment. This article will discuss the measures taken for analyzing heavy cargo exceeding thestandard weight limits. Also, a case study of a super heavyweight cargo weighing 548 tons
transporting an electric transformer over a 3-span bridge, designed in the early 70s, was carriedout. The bridge analysis included a documented deflection of a live load test before the actual
transportation. The results were conveyed into a FEM model of the bridge. Further precautions
were considered by monitoring the deflections of the bridge at the real-time transition of the
cargo and measuring the natural frequencies of the bridge before and after the transition.
1. Introduction
As the 21st century progresses, industries continually evolve and demand heavy cargotransportation. Consequently, trailer trucks are becoming more advanced, which places animmense strain on the infrastructure of roadways and bridges. Superload trucks are designed totransport heavy components such as bridge girders and massive power transformers. These
trailers have a particular configuration to evenly distribute the heavy load to multiple axles,
ensuring load distribution comparable to the statistical live load defined in the code. Concerning
the above, this article will discuss the following:a) Raise awareness about heavy loads' immediate and long-term effects on bridges.
b) Discuss bridge loading according to the design philosophy outlined in the codes.
c) Briefly discuss the dynamic amplification factor of heavy loads.
d) Present a case study, a 548-ton cargo transfer crossing over a 3-span girder bridge
2. Loading Definition
Loading is a crucial factor in the design of bridges. The selection of the primary design load isbased on identifying a single heavy vehicle, considering the probability of occurrence of multiple

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vehicle loads. National design codes commonly use three types of statistical loads to simulate
various loads, which are as follows:a) A system of concentrated loads with defined loads and distances.
b) An evenly distributed load across the traffic lanes, combining concentrated loads along thelanes.
c) An evenly distributed load across the traffic lanes, with less loading on some lanes.
The limit state design method is currently used in bridge design, with two essential
characteristics. Firstly, it considers all possible limit states, and secondly, it is based on
probabilistic methods.
The design depends on two parameters: the magnitude of the loads as it impinges the
structure – meaning the load effect, and the resistance or strength of the component.
Both the magnitude of the load effect and the resistance may be subjected to statistical variation,
but also according to the structure's life expectancy. Therefore, safety factors for each of them
are specified differently per country.
Extra attention should be paid when older bridges with outdated codes are subjected toheavy loads. Firstly, the bridge's condition may be unknown as it has been used for a substantial
amount of its expected lifespan. Secondly, the evolution of design codes tends to increase load
effects as industries develop and vehicles, such as trucks, become heavier, leading to larger
statistical loads being incorporated into the codes. Fig. 1 presents a schematic graph ofprobability distribution to illustrate the aging of a structure effect as discussed above. Over time
chances of failure increase, this can be seen as the area trapped in the middle of the graph.

Fig. 1. Probability density distribution for load effect and resistance graph – presents the effectof structure age on its vulnerability.
3. Examining Bridges for Heavy-Weight Cargo
As previously mentioned, a bridge design is based on a probabilistic approach. Therefore, when
examining a specific load on a bridge, it should always be compared to the statistical load in thedesign codes, considering the dynamic effects of the load. The ability of a bridge to withstandheavy loads depends not only on the weight of the cargo but also on the axle distribution andspan length ratio of the bridge. Specifically, the influence line of the bridge should be considered
concerning the specific load. Overall, it is crucial to thoroughly examine a bridge before
subjecting it to heavy-weight cargo to ensure its safety and structural integrity. The following
section will discuss an examination of a specific cargo load.

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4. Case Study: 548-Ton Cargo Transfer Across a Girder Bridge
This case study focuses on a three-span girder bridge on RD. 4 Israel, which crosses the Sorek
River. The bridge is composed of AASHTO III girders and a 20 cm cast-in-situ deck on top of
the precast beams and has a skew of 19 degrees. The spans have no continuity between them, so
each span acts independently as a simply supported beam; each span is about 1750 cm long. The
bridge was designed in the early 1970s and underwent rehabilitation in 2012. It needed to bechecked for a truck carrying a trailer with a 276-ton electric transformer, while the entire cargomass weighed 548 tons. The inspection and analysis of the bridge before the heavy-weight cargotransfer were crucial to ensure the bridge could safely handle the load. This case study highlights
the importance of regularly assessing and maintaining aging bridges to ensure their structural
integrity and safety.

Fig. 2. Cargo schema – dimensions and weight distribution
The verification process for whether the cargo can cross the bridge has two phases. Firstly, thecargo load is compared to the loads defined by the local country design code, which in Israel is
IS 1227, relying on BS 5400, specifically HA, HB, and HC. A Finite Element Method (FEM)
model of the entire structure was made based on the construction plan to ensure the mostaccurate simulation of the bridge's behavior. The FEM model in this case, is a grillage composedof longitudinal and transverse members simulating the deck. The model is then loaded with thecargo's axle distribution and the design code loads; this is done by defining a moving loadsystem to create an envelope of the maximum load effect. In order to consider uncertainties, afactor of 1.05 is provided for the cargo load; a dynamic factor could be neglected under thecondition that the cargo shall not exceed the driving speed of 5 kph. The bridge's criticalcomponents, i.e., longitudinal beams and transverse beams, are then examined for shear forces(in ULS) and bending moments (in SLS for prestressed members).
A second phase is required if the heavy cargo load exceeds the load effect of the designcode loads. In this case, the AASHTO girders of the bridge - the critical component - areexamined for stresses magnitude in the top and bottom fibers of the section, considering all
factors such as the construction sequence, dead load, superimposed dead load, temperature, Etc.
In the case of the cargo in question, a stress check was required since the cargo load had ahigher magnitude than the loads defined in the design codes. After undergoing the second phaseof checking the bridge from a design point of view, the resulting values of stresses in theServiceability Limit State (SLS) were hardly within the limitations defined in the Israeli code forprestressed concrete - IS 466/3.

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As discussed in section 2, the probability of a bridge failing increases as it ages. Given the factthat the bridge was constructed in the early 1970s - there is an ongoing debate about the best
course of action, which involves several options:a) Diverting the cargo to an alternate route,b) Temporarily reinforcing the bridge using steel beams and hydraulic jacks for theduration of the transition period, orc) Conducting a live load test on the bridge to measure its actual deflections andincreasing the accuracy of the calculation and modeling. The above would help calibrate
the model and consider the actual stiffness distribution between the longitudinal girders.
These options were carefully evaluated in collaboration with the authorities and the owner of thebridge - the national transport infrastructure company, Netivei Israel. Rerouting the cargo's
course was impossible, so the decision came down to either option (b) or (c). It's important tonote that temporary bridge strengthening in Israel for a one-time transition is rarely done andmust be considered meticulously, as it can affect the static schema.
Considering that the stress values were close to the specified code values, a recalibrationof the analytical model through a live load test was deemed sufficient. This option was chosen as
it increased the certainty level of the calculations and modeling without requiring significant
structural changes to the bridge.
4.1 Conducting a Live Load Test on the bridge
The live load test aimed to determine the bridge's actual stiffness in both longitudinal andtransverse directions. The test would enable us to calibrate the FEM model of the bridge
accordingly. The above could also assist in tackling any "red flags" before the transition of theheavy cargo, for instance, radical deflections or cracks. An experimental load plan was devised,
specifying the exact location of the loaded truck and providing instructions on the measurements
to be taken, as well as the expected theoretical deflection. The test truck weighed 104 tons andhad a trailer, corresponding to the bridge span in length. The location of the test truck path isidentical to the future path of the heavy cargo.

Fig. 3. Location of live load test and the gauges used to measure the deflection of the beams
shown on a transverse section of the bridge.
For measurement purposes, gauges with an error tolerance of 0.01 mm were used. As shown in
Fig. 3, measurement gauges were installed in three beams, each with one gauge at its midspanand two at its edges, for a total of nine gauges. The bridge's actual deflection was lower than

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predicted by the analytical model, indicating that the longitudinal girders are stiffer than initiallyanticipated. The founding of the measurements can be seen in table 1 below. Consequently, arevised recalibrated model was established, taking into account the stiffer beams. This resulted ina slightly higher bending moment per girder, meaning the transverse distribution between thegirders was reduced from the previous model. Even with the slightly higher values, re-examining
the calculation gave a satisfying result within the design codes limits, only with a higher amountof certainty.
Table 1. Theoretical and actual deflection table (mm)Testedbeam
Beam 3
Beam 4Beam 5

Theoreticaldeflection
(mm)
2.538
3.017
2.155
2.446
2.922
2.096
1.469
1.766
1.211

Test
point
MP-1
MP-2
MP-3
MP-4
MP-5
MP-6
MP-7
MP-8
MP-9

Actual deflection
(mm)
2.42
2.62
2.01
2.07
2.38
1.83
1.1
1.35
0.81

Deviation
(%)
4.9
15.2
7.2
18.2
22.8
14.5
33.5
30.8
49.5

Average
Deviation
(%)
9.1
18.5
38.0

4.2 Additional measurements during and after the crossing of the cargo
Verifying that the bridge would stay intact after the crossing of the cargo, additional measures
were taken. With the help of experts specialized in dynamic measuring equipment, the following
was carried out:•••

Data of the natural frequencies before the crossing of the cargo.
Deflection during the cargo crossing as a function of the time.
Data of the natural frequencies after the crossing of the cargo.

The deflection of the beams as a function of the cargo time traveling can be shown in Fig 4.
Consistently, beam 3 gets a higher deflection than beam 4; this difference can also be seen in thelive load test. Compering to deflections to theoretical calibrated model, the actual deflections arein fact lower, assuring the bridge state is on the safe side. Overall, changes in the natural
frequency were documented to be minor.

c

(a)

(b)

Fig. 4. (a) Beam 3 and 4 deflections (mm) as a function of time (sec) of the southern span of thebridge. (b) Maximum deflection from the FEM model (mm)147

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5. Conclusions
Running a probability-based analysis, the parameters involved in defining a limit state have an
uncertainty factor taken into account; these factors vary depending on the age and current state ofeach particular bridge. Accordingly, transporting heavy-load cargo on bridges should becautiously approached, mainly when dealing with older bridges. Conducting a live load test on
the bridge can provide a more precise estimate of its stiffness, enabling a more accurate
distribution of bending moments among its components. Adjusting the FEM model based on the
stiffness distribution between the longitudinal girders allows a more precise prediction of thebridge's behavior under various loads. Ultimately, this will aid in guaranteeing the bridge's safety
and longevity.
6. References
BS 5400 part 2 – specification for loads, 1978.
Dr. M. Mogilevski – Sorek bridge dynamic report, 2022.
System labs LTD – Sorek bridge live load test results report, 2022.
O'Connor, C., 2000. Bridge loads: An international perspective. CRC Press.
Israeli standard 1227 part 1 – loads on bridges: highway bridges, 1988.
Scholten, C., Enevoldsen, I., Arnbjerg-Nielsen, T., Randrup-Thompsen, S., Sloth, M., Engelund,
S. and Faber, M., 2004. Reliability-Based Classification of the Load Carrying Capacity of
Existing Bridges-Guideline Document (Report 291).
BSI, U., 2008. National annex to eurocode 1: Actions on structures–part 2: Traffic loads on
bridges, NA to BS EN 1991-2: 2003. British Standards Institution.

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Development of Pre-Tension Method for On-Site Production of Prestressed
Concrete Girders with Span Length between 20 m to 50 m
Dong-Woo Seo1*, Sangki Park2, Ki-Tae Park3, Hyun-Ock Jang4, and Yeon-Woo Shin5
1: Korea Institute of Civil Engineering and Building Technology, Goyang, South Korea; email:dwseo@kict.re.kr
2: Korea Institute of Civil Engineering and Building Technology, Goyang, South Korea; email:skpark@kict.kre.kr
3: Korea Institute of Civil Engineering and Building Technology, Goyang, South Korea; email:ktpark@kict.kre.kr
4: JEnC, Suwon, South Korea; email: hojang@jenc.kr
5: Taesung ENC, Sejong, South Korea; email: yw817@taesungenc.kr*: corresponding author

Keywords: Prestressed concrete girder; pre-tension; U-shaped; on-site production
Abstract: In Korea, the development of U-shaped prestressed girders is being attempted toincrease the length of I-type girders. However, since the dead weight due to the post-tension
method is large, a length of 30 m or less is common. In this study, the pre-tension method wasapplied without limiting the post-tension method to induce a decrease in self-weight andmaterials used due to a decrease in cross-sectional area. In addition, the author proposed theapplication of the in-situ pretension method using the internal reaction arm of the U-shaped
girder. The prestressed concrete U-shaped girder bridge is composed of a concrete deck slab anda composite section. Compared to prestressed concrete I-type girders, structural performance
characteristics such as resistance and stiffness are improved. Construction safety is also
improved during the fabrication and installation phases, and elongation is reduced due to girderweight reduction. Therefore, it is possible to secure the aesthetics and economic feasibility of thebridge. Accordingly, it is expected that efficient construction will be possible with high-quality
factory-manufactured and cast-in-place members. In this study, the pretension method wasintroduced to the field and the analytical performance of the anchoring block used for tension
was verified.
1. Introduction
Among the superstructure types, prestressed concrete I-type girders have gradually increased
their share due to improvements in design and construction methods over the past decade. As
such, the rigid frame superstructure type and the prestressed I-type girder used in road bridges
form a very large market and a lot of related technologies are being developed.
The advantage of rigid frame bridges is that they do not require repairs, such as no separate
movement or bridge support. However, its application in long-span bridges is limited. In addition,
prestressed concrete I-type girders have been introduced in Korea not long ago, so they havecomplex structures, ambiguous design standards, difficulty in construction (overturning due to149

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cross-sectional shape), and low economic feasibility. Feasibility due to increased material costs(Koo et al. 2005; Kim and Yang, 2008; Bae et al. 2017; Park et al. 2022).
Guo et al. A fully composite beam composed of U-shaped steel girders and angle connectors wasintroduced. They experimentally evaluated the beam and showed that it improved the bending
behavior (Guo and Liu, 2018). Svoboda et al. reinforcing and repairing a U-shaped reinforced
concrete bridge using post-tensioning with cable ducts. They concluded that post-tensioning withcable ducts is a very efficient way to reinforce existing bridges to increase their load-bearing
capacity (Svoboda et al. 2019). Jang et al. It is mentioned that the adaptability of U-shaped
girders is widely recognized due to the difference between U-shaped girders and conventional
section girders (Zhang et al. 2019).

(b)

(a)

Fig. 1. Configuration and conceptual diagram of the pre-tensioned U-type girder for the (a) unit
girder (precast and cast-in-place) manufacturing method and (b) closed bridge section
The PSC U-shaped cross-section girder bridge is composed of a concrete slab and a composite
section. It is a very efficient and economical bridge that increases safety at the construction stage,
lowers the height ratio by reducing dead weight, and secures the aesthetics of the bridge. It canalso improve the construction quality and efficiency of high-quality factory-made members andin-situ integral castings. Fig. 2 shows (a) the process of applying tension using an anchoring
block for tension and an internal reaction bar, and (b) the process of manufacturing a U-girder bypouring and curing concrete under tension.

(a)

(b)

Fig. 2. Pre-tension method using an internal reaction arm and anchor block for (a) Anchor blockhydraulic cylinder-internal reaction arm (b) Concrete pouring and curing
2. Pre-Tensioning Method
As shown in Fig. 3, in order to fix the tension between the lower slab of the U-shaped girder andthe outer wire, a tension fixing block is placed on both ends of the girder, and the outer wire150

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coming down from the inner deviator block pressurizes the hydraulic cylinder. It serves totighten the strands without deformation by external force.
Figure 3 shows a schematic diagram of the force-reaction concept of an anchor block using ahydraulic cylinder and multiple stranded wires. Although this anchor block was analyzed anddesigned as a compression member, it can show bending behavior if the thickness is thin, andtherefore it can be said that it is highly dependent on the strength of concrete and the thickness ofsteel plate.

Fig. 3. Force transmission mechanism between internal reaction arm-hydraulic cylinder andstrand anchorage
Figure 4 is an internal reaction arm (middle part of Fig. 4) using three PHC (Pre-stressed Highstrength Concrete) piles in which hydraulic cylinders are arranged, a cross section of a pretension U-shaped girder, and introducing an external twisted wire and pre-tension. Anchoring
blocks for tension are placed in rectangular positions. As long as the width of the girder does notchange, this anchor block is designed to be applied to all heights and may be slightly changed
depending on the cutting method of the strand in the future.

Fig. 4. Cross section of pre-tension U-type girder and anchorage block for tension
3. Consideration for Anchor Block for Pre-Tensioning
It must resist the tension of the outer strands for equilibrium of the inner reaction arm along withthe rigid composite anchor block for tension. Therefore, it is necessary to review the following
items before applying tensile force. Anchor blocks that can be moved and transported: A pair151

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must be able to move together as a set, the width and weight are limited for road movement, andthe crane capacity is easy to install on site.
It is expected that the application of UHPC as a fixing device for compression members will
greatly depend on the compressive strength of concrete by appropriately using it in considerationof usability such as weight and volume of ultra high strength concrete (UHPC). In the case of
UHPC using steel fiber, it is judged that the tensile strength of about 10 MPa is a spherical shape
and will be of great help to the tensile and shear resistance of the anchor block.

Fig. 5. Arrangement of internal reaction arm (PHC piles) and an anchoring block
A pretension method using a hydraulic cylinder that can be manufactured on site and is easy tomove for repeated production and a concrete pile used as an internal reaction force is proposed asshown in Fig. 5. A performance verification experiment of a full-size girder (L=40m, h=800mm)
is scheduled to be conducted in the near future.
4. Acknowledgments
This study was supported by the Ministry of Science and ICT (MSIT) grant by the Korea
Institute of Civil Engineering and Building Technology (KICT). (Project Number: KICT20230107).
5. References
Koo, M.-S., Kim, H.-H., Jung, Y.-D. 2005. A study of continuous PSC bridge with areinforcement steel plate, In Proceedings of the Computational Structural Engineering Instituteof Korea Annual Conference (Spring), 422–429.
Kim, K.-S., Yang, I.-H. 2008. PSC development of new type PSC beam girder bridges in Korea.
Magazine of the Korea Concrete Institute, 20, 26–33.
Bae, K.-M., Min, K.-H., Lee, C.-O., Lim, N.-H. 2017. Analysis and improvement of long-term
deflection of PSC I girder for railway bridges. Journal of Korean Society of Hazard Mitigation,
17, 9–15.
Park, S., Kim, J., Jung, K.-S., Seo, D.-W., Park, K.-T., Jang, H.-O. 2022. On-site construction
method for U-girder with pre-tension and verification of analytical performance of anchoring
block. Journal of Korea Society of Disaster & Security, 15(3), 67-77.

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Guo, L., Liu, Y.,Qu, B. 2018. Fully composite beams with U-shaped steel girders: Full-scale
tests, computer simulations, and simplified analysis models. Engineering Structures, 177, 724–
738.
Svoboda, A., Klusáček, L., Olšák, M. 2019. Strengthening and rehabilitation of U-shaped RC
bridges using substitute cable ducts. Advances in Materials Science and Engineer, 1-21
Zhang, J., Jing, Y., Li, P., Han, W., Zhang, N., Zhou, Y. 2019. Experimental and numericalinvestigation on the ultimate vertical bearing capacity of U-shaped girder with damaged web.
Sensors, 19, 3735.

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Environmental Effects

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Temperature-Induced Variation in Geometric Characteristics of Existing Cracks in
Double-Box Girders of Concrete Cable-Stayed Bridge
Xujia Liu 1,2, Youliang Ding 1,2*, Hanwei Zhao 1,2, Chunfeng Wan 1,2, and Fangfang Geng3
1: Key Laboratory of Concrete and Pre-stressed Concrete Structures of the Ministry of Education,
Southeast University, Nanjing 210096, China
2: College of Civil Engineering, Southeast University, Nanjing 210096, China
3: School of Architecture Engineering, Nanjing Institute of Technology, Nanjing 211167, China*: Corresponding author
Keywords: crack propagation, service life, concrete cable-stayed bridge, temperature field
Abstract: The crack propagation of concrete cable-stayed bridge’ main girder directly affects theservice life of the bridge. To explore the influence of the temperature field on the existing crack
width of the double-box main girder, a long-span concrete cable-stayed bridge is taken as an
example. The temperature field and crack width of the main girder of concrete cable-stayed
bridge are monitored. The analysis of the measured data shows that the variation of the innersurface temperature of the box in winter, spring and summer is approximately a cosine curve
trend, and the maximum and minimum temperature in the main beam box are 35 ℃ and 0 ℃ insummer and winter respectively. The cracks in the inner wall of the box section of the main beam
have obvious closing and opening phenomena with the change of temperature, and difference
between maximum and minimum width approximately 0.9 mm, and the crack opening andclosing trend shows a sinusoidal curve with the season, which is opposite to the temperaturechange trend, and the temperature-crack width shows a monotone decreasing linear correlation.
It is difficult to repair cracks in the main girder due to the phenomenon of crack closure andopening, and the effect of conventional repair methods is not ideal, which is very consistent withthe phenomenon of cracking again after crack repair.
1. Introduction
Due to the material characteristics of concrete, initial cracks will inevitably occur in concrete
cable-stayed bridges during construction. After the concrete cable-stayed bridge enters theoperation period, the crack width further increases due to the adverse factors such as vehicle andtemperature. After the crack reaches a certain width, the harmful substances will go deep into theconcrete, which will further lead to steel bar corrosion, concrete carbonization and otherphenomena, at the same time, the crack may further expand. The sunshine temperature effect is
an important factor in the process of crack formation and expansion. Therefore, in view of theexisting cracks in concrete cable-stayed bridges, it is very necessary to study their temperature

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variation characteristics.
In order to study the temperature variation characteristics of existing cracks in concrete
cable-stayed bridges, we should focus on the analysis of the temperature field characteristics ofthe bridge. On-site monitoring is an effective way to study the thermal field characteristics of thebridge. Many researchers have studied the temperature field of the bridge. For example, in early
1983, the temperature distribution of the composite box girder bridge was studied according to
the monitoring data of the Maskova River Bridge (Dilger et al., 1983). It is concluded that thetemperature difference is affected by geometric characteristics, material types and environmental
factors. In order to study the temperature field distribution characteristics of long-span flat steel
box girder suspension bridge, the statistical characteristics of temperature field and predicted theextreme temperature difference by using the long-term field measurement data of Runyang
suspension Bridge in China was analyzed (Ding et al., 2012). Taking Incheon Bridge in South
Korea as an example, the distribution law of temperature field of steel box girder of cable-stayed
bridge by numerical method was analyzed (KIM et al., 2015).
In order to further analyze the temperature variation and distribution characteristics of thebridge, as well as the temperature variation characteristics of the existing cracks. This paper
takes a concrete cable-stayed bridge in China as the research background, based on themonitoring system of concrete cable-stayed bridge. The long-term temperature monitoring dataand crack width time-varying data are obtained, the correlation characteristics between bridge
temperature and crack width are analyzed, and the temperature-crack width fitting curve of
concrete cable-stayed bridge is established. so as to provide a basis for bridge health and safety
design, management and maintenance.
2. Example
The bridge has a total length of 2062 m, a span of 152 m + 370 m + 152 m, a total width of 38.6
m and six lanes in both directions, as shown in Fig. 1 (a). Construction of the bridge began in
March 2003, closed in August 2005, and was completed and opened to traffic at the end of 2005.
The main girder is prestressed concrete box girder. The standard section of the main beam adopts
a double-side box section, with a height of 3.2 m, a roof thickness of 0.3 m, a bottom plate
thickness of 0.4 m, an edge oblique web thickness of 0.3 m and a middle web thickness of 0.4 m,
as shown in Fig. 1 (b). The non-standard cross section of the main beam (cable tower position
main beam section, side span cable dense area main beam section, side span closure section main
beam section) adopts single box and four chamber section. Longitudinal and horizontal limiting
devices are arranged at the junction of the tower and beam. The bridge cloth is equipped withtemperature sensor and crack sensor, and the data acquisition frequency is 1/300 Hz. The specificlocation of the sensor is shown in Fig. 1 (a), (b), (c) and (d). (BTEM/ temperature sensor, CRW/
crack sensor)

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(a) Bridge elevation(b) Cross section(c) Sensor placement point at the location of the 27# girder(d) Sensor placement point in the middle of the span
Fig. 1. Bridge structure and layout of measuring points
3. Analysis
As can be seen from Fig. 2 (a), from November 25, 2021 to July 22, 2022, the lowest andmaximum temperature inside the main beam of the 27# bridge tower is 3.3 ºC and 34 ºC, thetemperature difference between winter and summer is 30.7 ºC, the minimum and maximum
temperature in the middle of the span is 0.4 ºC and 39.3 ºC, and the temperature difference
between winter and summer is 38.9 ºC. As can be seen from Fig. 2 (b), from November 25, 2021
to July 22, 2022, the minimum and maximum values of roof crack width inside the main girderof No. 27 bridge tower are -0.735 mm and 0.278 mm, and the difference is 1.013 mm. The
minimum and maximum values of web crack width in the middle of span are -0.428 mm and0.278 mm, and the difference is 0.706 mm. Compared with Fig. 2 (a) and (b), the temperatureand crack width show reverse change, that is, from November 25, 2021 to February 17, 2022, thetemperature is lower and the crack opening is more obvious, while from June 9, 2022 to July 22,2022, the temperature is higher and the crack shows a closing trend.

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(a)

(b)

Fig. 2. Long-term monitoring data of temperature and crack width: (a) BTEM, (b) CRW
Fig. 3 shows the correlation between temperature and crack width (the monitoring data of
temperature and crack width are averaged with 2 hours as the section). From Fig. 3 (a) and (b), it
can be seen that the crack width decreases linearly with the increase of temperature. Through themeans of data fitting, it can be seen that the crack opening is especially obvious when theinternal temperature of the 27# main beam is close to 0 ºC, and shows the same characteristics inthe middle of the span. According to the fitting curve of Fig. 3 (a) and (b), the temperaturechange of the main beam in winter and summer is obvious, but the opening and closing of theroof crack in the 27# main beam is obvious, and the opening and closing of the roof crack ismore sensitive than that in the web.

(a)

(b)

Fig. 3. Temperature-crack width correlation: (a) BTEM/CRW-1, (b) BTEM/CRW-2

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4. Conclusion
The temperature field of concrete cable-stayed bridge is obviously related to the opening andclosing degree of cracks, showing a linear decreasing trend as a whole, that is, the higher thetemperature is, the smaller the crack width is. When the internal temperature of the bridge is low,
the internal cracks of the bridge are in the trend of opening and expanding. At the same time, thevariation trend of roof cracks in the box section of the bridge with temperature is more obvious
than that of the web.
5. References
Dilger, W., Ghali, A., Chan, M., Cheung, M., and Maes, M.. 1983. Temperature stresses incomposite box girder bridges, Journal of Structural Engineering, ASCE, 109(6), 1460-1478.
Ding, Y. L., Zhou, G. D., Li, A. Q., and Wang, G. X.. 2012. Thermal field characteristic analysisof steel box girder based on long-term measurement data, International Journal of Steel
Structures, 12(2), 219-232.
Kim S. H., Park S. J., Wu J. X.. 2015. Temperature variation in steel box girders of cable-stayed
bridge during construction, Journal of Constructional Steel Research, 122, 80-92.

159Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Utilizing Pressurized Injection of Lithium Nitrite in Treating Alkali-Silica
Reaction and Salt Damage in Concrete Structures
Shohei Notsu1*, Tatsuya Kitada2, Son Van Nguyen3, and Kazunori Era4
1

: Kyokuto Kowa Corporation, Hiroshima, Japan; email: notsu@kkn.co.jp: Kyokuto Kowa Corporation, Hiroshima, Japan; email: t-kitada@kkn.co.jp
3: Kyokuto Kowa Corporation, Hiroshima, Japan; email: n-son@kkn.co.jp
4: Kyokuto Kowa Corporation, Hiroshima, Japan; email: era@kkn.co.jp
2*: corresponding author

Keywords: Alkali-silica reaction (ASR); Chloride contamination; Pressurized injection; Repair
Abstract: In recent years, a technique using nitrite ions has attracted attention as a means ofrepairing concrete that has been deteriorated due to salt damage. This paper verifies theapplicability of the pressurized lithium nitrite injection method to salt damage, which wasintroduced as an Alkali-Silica Reaction (ASR) repair technology by (Kitada & Era 2019).
Additionally, an example of application of this method to a bridge pier that has been deteriorated
due to both ASR and salt damage is presented and the results are discussed.
1. Introduction
Concrete is a porous material that absorbs the salt used for de-icing the snow during winter season
and airborne sea salt, which causes damage and deteriorates concrete. To repair the concrete thathas been deteriorated due to salt damage, crack injection or surface protection is often used toblock the chloride ions intrusion from the outside. However, even if the chloride ions intrusionfrom the outside is prevented, the corrosion will still continue to progress if the passive film of therebar has already been damaged. Since the corrosion environment cannot be improved internally,
re-deterioration continues to occur. Therefore, recently, attention is given to a repair technique that
focuses on the regeneration of passive film on reinforcing bars by using nitrite ions (Era et al.
2010).
To apply the pressurized lithium nitrite injection method, a small hole is drilled in the affectedconcrete member, and lithium nitrite is injected into the concrete with a high pressure. (Kitada &Era, 2019) have explained this method in details.
A schematic diagram of this method is illustrated
in Fig. 1.
Lithium nitrite used in the pressurized lithium
nitrite injection method is a repair material
consisting of nitrite and lithium ions. (Kitada &Era, 2019) focused on the effect of nitrite ions on
suppressing the expansion of reactive
aggregates, and introduced the pressurized
lithium nitrite injection method as a repair
technology for concrete deteriorated by AlkaliSilica Reactions (ASR).
Fig. 1. Schematic diagram of the method160

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Additionally, nitrite ions have the effect of regenerating the passive film and improving thecorrosion environment of the rebars. Thus, they can be suitable repair material to be used inrepairing salt damage. This paper verifies the effect of applying pressurized lithium nitrite injection
method to salt damage, as well as presenting example of applying this method to a bridge pier thathas been deteriorated due to both ASR and salt damage.
2. Verification of Pressurized Lithium Nitrite Injection Method on Salt Damage
2.1 Experiment outline
G0
In order to verify the anticorrosive property of pressurized
G1
lithium nitrite injection method on rebars, (Kojima et al 2020)
G2
created a specimen simulating reinforced concrete
deteriorated by salt damage and conducted a laboratory Fig. 2. Rebar inside specimen
experiment.
Two specimens (150 x 150 x 150 mm, lithium nitrite injected
and 100 x 100 x 150 mm, with no lithium nitrite) wereconstructed using ordinary Portland cement, with a water
cement ratio of 55%. Three round rebars with a dimension of
φ10×130mm were placed with a 40 mm concrete cover ineach specimen. In addition, in order to simulate deterioration
due to salt damage, Cl-: 5.0 kg/m3 was mixed in the concretein advance. The embedded rebars were non-corroded rebars
G0 and corroded rebars with different degrees of corrosion(G1 and G2) as shown in Fig. 2.
The specimens were air cured for 28 days, and on 67th day,an 80 mm hole was drilled in the center of the specimen as
shown in Fig. 3. Lithium nitrite was injected with pressure
ranging from 0.3 to 0.5 MPa. The injection was continued
Fig. 3. Experiment summary
until the NO2-/Cl- molar ratio in the concrete reached 1.0 for
the premixed Cl- amount of 5.0kg/m3. The injection was done for a maximum of 7 days.
Hereinafter, the injected specimen will be referred to as M1 and the specimen which was notinjected with lithium nitrite will be referred to as M0.
After injection was completed, the specimens were placed in a wet environment with a temperatureof 20°C and a humidity rate of 90% for 190 days, during which the self-potential was measured.
After that, the surface concrete of the specimen was cut with a concrete cutter to reduce theconcrete cover of the specimen to 20 mm and the specimens were continuously exposed to an
environment which simulated a salt-damaged environment while self-potential and Polarization
resistance was being measured periodically. 340 days from the start of salt water penetration, adismantling test was conducted to confirm the corrosion status of rebars, and to measure thecorrosion weight, corrosion area ratio, and nitrite and chloride ion amounts. Next, the results ofself-potential and polarization resistance are presented.
2.2 Results and discussion
Fig. 4. shows the time period of spontaneous potential for M0 and M1 specimens. The selfpotential of the M0 specimen without pressurized lithium nitrite reduced immediately after it wasleft in a wet environment, and corrosion progressed early. On the other hand, the self-potential of
the M1 specimen which was injected with lithium nitrite was maintained during the wet

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G0M0
G0M1

Self-potential (mV) VS SCE

-200

G1M0
G1M1

G2M0
G2M1

G0M0
G0M1

50

-300

G1M0
G1M1

G2M0
G2M1

40

-400

30

-500
-600

20

-700
-800

Polarization resistance（kΩ･㎠）

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

10

0

100

200

DrynessDampness
Injection

300
Period (day)

400

500

Brine penetration600
dismantlingFig. 4. Change overtime of self-potential

0

0

50

100 150 200 250 300
Brine penetration period (day)350

Fig. 5. Change in polarization resistanceenvironment period. However, the potential of the specimen with G2 rebar showed a reduction
during the period of salt water infiltration. The self-potential at the time of dismantling was about
the same as the specimen without injection, suggesting that corrosion had progressed. This isprobably because the NO2-/Cl- molar ratio decreased with the penetration of Cl-. Therefore, in anenvironment where there is a chance of new Cl- penetration, a better practice is assumed to beusing the pressurized lithium nitrite injection together with a surface protector material.
Fig. 5 shows the change in polarization resistance of G0, G1 and G2 rebars over a period of 350
days. G0 and G1 rebars always showed higher values in M1 than in M0. However, there was atendency for a gradual decrease during the measurement period.
On the other hand, the strength of G2 reinforcing bars was higher than that of M0 until about 150
days in M1, but decreased with the number of days immersed in salt water, until it had the samevalue as M0 after about 280 days of measurement.
As mentioned above, the long-term rust prevention effect of this method was confirmed from theself-potential and polarization resistance results. However, it has been suggested that the NO2-/Clmolar ratio in the concrete may decrease due to the penetration of Cl- after the injection, especiallywhen the corrosion of the rebar has progressed to the level of G2 rebar before the injection, andthe corrosion prevention effect may be impaired. Therefore, if the reinforcement corrosion issevere, it is considered necessary to suppress the intrusion of Cl- by combining surface protection
work together with pressurized lithium nitrite injection method.
3. Example of Practical Application to Actual Structures
3.1 Outline of T-bridge
To utilize the developed technology, we have applied the pressurized lithium nitrite injection
method to a T-bridge pier which has been deteriorated due to both ASR and salt damage, in 2017.
Exposed rebar
Pier6

Fig. 6. Appearance of T-bridge

162

Fig. 7. Deterioration of T-bridge

Bridge Engineering Institute Conference 2023 (BEI-2023)
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The T-Bridge is a road bridge that was constructed in 1965 and is located on the Seto Inland Sea
side of the Hyogo Prefecture, Japan as shown in Fig. 6. As it can be seen in Fig. 7, the P6 pier has
been deteriorated as the rebars are exposed and there are visible map cracks in the surface.
In addition, a high chloride ion concentration of 3.8 kg/m3 was detected, and according to theresults of residual expansion test as per ASTM C1260, it exceeded 0.4% after 28 days ofmeasurement, which is quite beyond the hazardous standard of 0.10%.
Based on the above investigation results, it was determined that the damage to the P6 pier of TBridge was due to combination of salt damage caused by salt blown from the sea, and ASR reaction
due to alkali expansive aggregate. The pressurized lithium nitrite injection method was adopted asa counter measure.
3.2 Confirmation of effects of the method
To confirm the effects of applying this method, samples were collected from three different
locations of the concrete member after the injection, and a residual expansion test was conductedusing the JCI-DD2 method (current JCI-S-011 method). The amount of residual expansion in all
samples was below the reference value of 0.05%. Therefore, it was determined that the pressurized
lithium nitrite normally permeated and diffused into the concrete, and that the supply of lithium
nitrite made the alkali silica gel non-swelling. In addition, visual follow-up surveys are being
conducted every few years after injection, and it has been confirmed that no re-deterioration has
occurred as of 2022, five years after applying the pressurized lithium nitrite injection.
4. Conclusion
1) It was confirmed that the pressurized lithium nitrite injection method has a long-term rustprevention effect on reinforced concrete deteriorated by salt damage. However, if Cl- penetrates
the concrete after applying the pressurized lithium nitrite injection method, the NO2-/Cl- molar
ratio will decrease and the rust prevention effect will be vanished. If rebar corrosion is severe,
it is necessary to combine this method with surface protection work to achieve desirable
outcome.
2) As a result of applying the pressurized lithium nitrite injection method to a bridge pier that havebeen deteriorated due to both ASR and salt damage, the amount of residual expansion due to
ASR was reduced. In addition, even as of 2022, five years after the repair, there were no signsof deformation or further deterioration.
5. References
Era, K, Tokuno, T, Minematsu, S, & Miyagawa, T (2010) “Study on the Long-term Durability ofLithium Ion Pressurized Injection, Proceedings of the concrete structure scenarios”, JSMS, Vol10,
pp. 167- 172,
Kitada, T, & Era K (2019) “Treatment of Alkali-Silica Reaction in Concrete Structures Using
Pressurized Injection of Lithium Nitrite”, Bridge Engineering Institute Conference 2019 (BEI2019), Honolulu, Hawaii, USA, July 22-25, 2019, pp. 36-40,
Kojima, A, Hazehara, H, Okabe, Y, & Era, K (2020) “Evaluation of long-term corrosion protection
performance of concrete with different amounts of Li nitrite pressurized into the concrete undersaline degradation environment.”, Proceedings of the Japan concrete institute, Vol.142, No.1, pp.
1468-1473 (in Japanese),163

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Fatigue of Steel Anchor Box on Cable-Stayed Bridges Considering Thermal
and Vortex-Induced Vibration under Random Traffic Flows
Zhiwen Zhu 1*, Shuang Yan2, Federico Accornero3, and Yichun Deng4
1: Shantou University, Shantou China; email: zhuzw@stu.edu.cn
2: Shantou University, Shantou China; email: syan@stu.edu.cn
3: Shantou University, Shantou China; email: federico@stu.edu.cn
4: Hunan University, Changsha China; email: 913241281@qq.com*: corresponding author
Keywords: fatigue; steel anchor box; VIV of stayed cable; random traffic; thermal effects
Abstract: In order to evaluate fatigue performance of the steel anchor box with its stay cable
presented with large amplitude vibration under in-service condition, continuous monitoring was
carried out on a long-span cable-stay bridge to measure the acceleration of the stay cable andstress at details of its steel anchor box (SAB). The characteristics of stay cable vibration, as wellas the stress at the SAB details contributed by loadings from vortex-induced vibration (VIV) ofstay cable, passing vehicles and thermal action were investigated in time- and frequency domains.
Hence the loading mechanism of VIV of stay cable and thermal action on the SAB wasinvestigated. Based on the nominal stress method, the fatigue performance of the SAB underjoint action of VIV of the stay cable, truck passage and thermal action were evaluated. It is foundthat the significant vibration of stay cable, characterized by the high-order multi-mode VIV anddominated by in-plane vibration with peak frequency mode appeared among the fifth to theseventeenth, occurred at the stay cable JB01 under mean wind speed range of 2m/s~9m/s, withobserved maximum in-plane peak acceleration of 25m/s2. The thermal action contributes
significantly to the maximum stress range at the details of the SAB, but it only creates one stresscycle each day. Compared to the thermal action, the stress range generated by the passage ofvehicles is low, but trucks produce significant large number of loading cycles; The inertial force
produced by the VIV of the stay cable is applied to the SAB, but only produces extremely low
stress at those details. Consequently, its effects on stress and fatigue of SAB are completely
negligible. It is concluded that the fatigue evaluation of the steel anchor box needs to consider
the thermal action. However, even considering the joint action of the VIV loading, thermal effectand truck passages, the fatigue life at the critical details of the SAB, i.e., the deck-side weld endof the upper and lower plate to the outer web, as well as both weld end of the bearing plate to theouter web of steel box girder, is remarkable higher than 100 years. Hence fatigue performance of
the SAB under in-service condition satisfies the bridge design requirement.
1. Introduction
The complicated details of the steel anchor box (SAB), and the large and concentrated force
acting by the stayed cable, may result in severe stress concentration and high local stress at somedetails. So far, only few studies on the fatigue performance of the SAB of cable-stayed bridge
have been reported (Lin et al, 2012; Wu, 2020; Zhu et al, 2021). For cable-stayed bridges underin-service condition, the repeated passing trucks on the bridge deck will produce fluctuated axial

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forces in the stayed cables, which will act on the SAB and will produce repeated fluctuatingstress at the details of the SAB. In case of vortex-induced vibration (VIV) of the stayed cable(Liu, 2021), the inertia force of the stayed cable (VIV force) apply to the SAB at the cable end
may be large, and the SAB will subject joint action resulting from the VIV force of the cable,
truck loading on deck and thermal action on the whole bridge. The joint of the three actions may
produce a high stress range in the welding details of the SAB, which may present fatigue issue atmain girder connection with the stayed cables.
2. Test Setup
The propeller anemometer is installed at the centre of the main span at height of 5m above bridge
deck, and a bidirectional accelerometer is fixed on the cable JB01 at the height of 6.22m. The
strain gauges were mainly applied at those locations with severe stress concentration on the outerweb. All sensors are connected to the data-acquisition system and then data are sent into themonitoring system. Figure 2 shows the arrangement of strain gauges applied at the fatigue-prone
detail of the SAB.

Fig. 1. Monitoring system
3. Vibration Monitoring and Analysis
Figure shows the vibration response of JB01 cable. It is clear that the observed cable vibration isdominated by its in-plane component, and the peak acceleration first rises and then drops with
the increase of wind speed. In addition, the participating modes distribute among the 5th to the
17th. It is concluded that the observed cable vibration is the high-order multi-mode VIV.
Deck plate
L

U1

Upper plate
U2

Bearing plate
Outer web1

Peak acceleration (m/s2)

Lower plate
Stiffeners

B1

Shim plate
Inclinedbottom plate

B4

B2
B3

1

25

20

20

15

15

10

10

5

5

0

0

0
5
10
15
Deck wind speed (m/s)

Diaphragm

Fig. 2. strain gaugearrangement25

0

5 10 15 20
Mode No.

(a)(b)
Fig. 3. Vibration response of JB01 cable, (a) time history; (b)peak acceleration

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2000

40

L

10

U1

t (s)

20

U2

30

Fig. 4. Typical stress records at detailsof the SAB under truck passage

(a)

0.5

1.5 1 1.5 1

1 0.5

Si (MPa)

44-46

2
0

0

26.5

20

30

40-42

B5

20

38-40

10

Σni=4238
Sr,max=45.0MPa

200

32-34

B4

B3

0

398.5

400

30

28-30

20

18-20

10

14-16

0

B2

6-8

4
0
-4
-8

4000 3806

B1

8-10

4
0
-4
-8

6000

ni

4
0
-4
-8

4-6

σ (MPa)σ (MPa)σ (MPa)4. Stress Monitoring on SAB
Figure provides typical stress curves at details of the SAB under passage of a same truck. Thereis only one stress cycle created at those locations. It suggests that the SAB behaviors as the mainloading-bearing component in the bridge, which only produce one stress cycle at its fatigue detailunder one truck passage.

(b)

Fig. 5. 7-day measured stress records and stress
spectrum at U1,(a) stress records; (b) stress spectrum

Figure plots the continuously measured stress and stress spectrum at locations U1 in seven days.
The seven-day stress curves show seven long-period waves, which is characterized by one cycle
a day, significantly high stress than truck loading and variable-amplitude stress ranges thathighly depends on the weather condition, such as the solar radiation, etc. If the thermal effect is
excluded, the maximum stress ranges from the measured stress is only 10.7MPa. In addition, forthe stress ranges higher than that in the seven-day measured stress spectrum shown in Figure 5,
there is only seven cycles. Hence every day the thermal effect created one stress cycle. Check onmonitored stress time histories at U2, L, B1, B1, B3 and B4 show similar features. Thissignificantly high stress range is resulted from thermal effect on the whole bridge, which resultsin fluctuating forces in the stayed cable. Compared the maximum stress with and without thermal
effect, thermal effect made major contribution to high stress ranges, hence it should beconsidered in fatigue evaluation of the SAB.
5. Loading Mechanism of VIV Force on SAB
When the bridge was in normal service condition without VIV of the stayed cable, the in-plane
acceleration time history of the stayed cable and its counterpart of stress time history at L on the
SAB are shown in Figure 6. Two periods of cable acceleration and stress at L are selected, i.e.,
one from 30s to 60s without truck loading, and another from 75s to 105s with a truck passing thebridge. The normalized spectrums of the two periods show no difference in frequency domain.
Hence the truck passage provides little contribution to the cable vibration, and provides minor
effect to the normalized spectrum of the stress time history.

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acceleration

0

30

60

No truckσ (MPa)

10
5
0
-5
-10

Truck

0

30

90

120

150
L

Truck

60

90
t (s)

180

120

150

180

1.0

acceleration

0.5
0.0
1.0

0

4

8

12

16

20
L

0.5
0.0

0

4

(a)

8

12

Frequency (Hz)16

20

Normalized spectrum Normalized spectrum

No truckσ (MPa)

0.4
0.2
0.0
-0.2
-0.4

Normalized spectrum Normalized spectrum

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023
1.0

acceleration

0.5
0.0
1.0

0

4

8

12

16

20
L

0.5
0.0

0

4

(b)

8
12
Frequency (Hz)

16

20

(c)

Fig. 6. Time- and frequency-domain plot of acceleration and stress when stayed cable without
VIV, (a) time history; (b) frequency-domain plot in 30s to 60s; (c) frequency-domain plot in 75s
to 105s

30

60

90

120

0

30

60

150

180
L

No truckσ (MPa)

10
5
0
-5
-10

acceleration

0

Truck

90
t (s)

(a)

120

150

180

1.0
f13=12.70Hz

0.5
0.0
1.0

acceleration

0

4

8

12

16

20
L

f13=12.70Hz

0.5
0.0

f14=13.67Hz

04

f14=13.67Hz

8
12Frequency (Hz)

(b)

16

20

Normalized spectrum Normalized spectrum

Truck

No truckσ (MPa)

24
12
0
-12
-24

Normalized spectrum Normalized spectrum

When the bridge was in normal service condition, but the VIV of the stayed cable occurred, thein-plane acceleration of the stayed cable and its counterpart of stress time history at L are shownin Figure 7. The VIV force is applied on the SAB, but its loading effect is quite smaller than thatof the truck loading. However, even under the truck loading, the normalized spectrum of stresstime history at L clearly indicates the peak frequency, which agrees well with the normalized
spectrum of cable acceleration. It confirms that the SAB undergoes joint action of truck loading
and VIV force, although the loading effect of the VIV force is significantly lower than the truck.
1.0
f13=12.70Hz

0.5
0.0
1.0

acceleration

0

4

8

16

20
L

0.5
0.0

12

f13=12.70Hz
0

4

8
12
Frequency (Hz)

16

20

(c)

Fig. 7. Time- and frequency-domain plot of acceleration and stress when stayed cable with VIV,(a) time history; (b) frequency-domain plot in 30s to 60s; (c) frequency-domain plot in 150s to
180sCompared to the thermal action and truck loading applied to the SAB, the VIV force produces
extremely low stress at the SAB, and the effect on stress and fatigue is completely negligible.
6. Fatigue Life Estimation of SAB
The SAB is subjected to the variable-amplitude loading by trucks, thermal effect, or joint actionof both, so the stress spectrum of the details should be obtained based on the statistical countingof the measured stress using the rain-flow counting method. Then calculated the fatigue life of
the details according to Miner’s law of linear cumulative damage and the S-N curve in AASHTOLRFD (AASHTO, 2017).

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Table 1 shows the fatigue evaluation results at various locations of SAB. The maximum stressranges at U1, L, B1 and B2, are all higher than the CAFL of category E, as a result, finite life isexpected at the four locations. The estimated fatigue life is significantly higher than the bridge
design life of 100 years. However, the measured maximum stress ranges at U2, B3, B4 and B5 arelow. Since their fatigue category is higher than category E, their fatigue life is infinite evenconsidering the thermal effect, hence fatigue will not be a concern at those locations.
Table 1. Fatigue evaluation results at various locations of SAB
Maximum stress range (MPa)Fatigue life (year)NO.
measured
without thermal effectmeasured
without thermal effect
U1
45.0
10.7
2740.3
infinite
U2
24.5
6.4
infinite
infiniteL
33.0
7.1
19020.7
infinite
B1
47.7
5.3
8190.8
infinite
B2
31.5
3.6
50369.2
infinite
B3
28.1
7.0
infinite
infinite
B4
21.0
6.8
infinite
infinite
B5
33.5
5.4
infinite
infinite

CAFL
(MPa)
17.9
69.0
17.9
17.9
17.9
69.0
69.0
69.0

7. Conclusions
Measurement of the cable acceleration and stress at details of the SAB was carried out for thestayed cable JB01 on a long-span cable-stay bridge. The feature of cable vibration wasinvestigated, and the fatigue performance of the SAB under joint action of the VIV force of thestayed cable, truck passage and thermal action was evaluated, which will provide goodsuggestions for the fatigue research and anti-fatigue design of SAB.
8. Acknowledgements
Financial support for this study was provided by National Natural Science Foundation of China(51878269, 52278509), to which the writers gratefully appreciate.
9. References
AASHTO (2017). AASHTO LRFD Bridge Design Specifications. 8th Ed., American Associationof State Highway and Transportation Officials, Washington D.C.
Liu, Z.W., Shen, J.S., Chen, Z.Q., et al (2021). Experimental study on aerodynamic control
measures for vortex-induced vibration of stay-cable. Journal of Vibration Engineering, 34(03)441-451. (In Chinese)
Lin, C. H., Lin, K.C., Tsai, K.C., et al. (2012). Full-scale fatigue tests of a cable-to-orthotropic
bridge deck connection, Journal of Constructional Steel Research, 70(3), 264-272.
Wu, Q., Zhu, Z.W., Li, J.P. (2020). Fatigue evaluation of connection details on cable anchorage
based on field measured stress spectrum. Journal of Railway Science and Engineering, 17(11)2823-2831. (In Chinese)
Zhu, Z.W., Li, J.P., Cai, J.Y., et al. (2021). Field investigation on fatigue of steel anchor box in168

Bridge Engineering Institute Conference 2023 (BEI-2023)
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cable-stayed bridge under in-service condition considering vortex-induced cable vibration,
Journal of Vibration Engineering, 34(1), 9-19. (In Chinese)169

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Mechanism of Control Method for the Alkali-Silica Reaction and Delayed Ettringite
Formation Expansion Using Artificial Lightweight Aggregate
Atsushi Teramoto1* and Moka Tanabe2
12: Hiroshima University, Higashi-Hiroshima, Japan; email: a-teramoto@hiroshima-u.ac.jp: Hiroshima University, Higashi-Hiroshima, Japan; email: m221023@hiroshima-u.ac.jp*: corresponding author

Keywords: ASR; DEF; artificial lightweight aggregate; X-ray CT
Abstract: The alkali–silica reaction (ASR) and delayed ettringite formation (DEF), which are
typical expansive deteriorations of concrete structures, can be reduced by mixing artificial
lightweight aggregate (ALA) with concrete. In this study, to clarify the mechanism by which
ALA reduces ASR and DEF expansion, artificial lightweight fine aggregate (ALFA) andartificial lightweight coarse aggregate (ALCA) were mixed into specimens with accelerated
deterioration. Thereafter, length change measurements and X-ray computed tomography (CT)imaging were performed.As a result, the degree of ASR expansion decreased as the proportion of
ALA increased, and this effect was particularly significant for ALFA. With respect to the ASR,
ALFA replaced 50%, and for DEF, a 100% replacement of the normal fine aggregate almost
completely suppressed the expansion. The internal cracking of each specimen was observed by Xray CT imaging, and the degree of cracking decreased in the specimens with reduced expansion,
thus suggesting that the driving force of the expansion was reduced. Furthermore, with respect to
the ASR, the amount of dissolved silica and decrease in alkali concentration were measured
over time for ALFA and ALCA. It was confirmed that the fine powder content of ALA
contributed particularly to alkali adsorption. These results confirm that ALA reacts with the alkaliin the concrete and reduces its concentration. Finally, we conducted a numerical simulation toestimate the degree of expansion by uniaxial reactive model using various coefficientsdetermined based on the results of aggregate reactivity tests. The results revealed that the degreeof ASR expansion could be estimated by considering the alkali reactivity of ALA. Moreover, themechanism of DEF reduction could not be numerically explained. As a hypothesis, the reaction of
ALA powder with calcium hydroxide in the concrete via the pozzolanic reaction may reduce thepotential of pH in the concrete system, thus resulting in the suppression of DEF.
1. Introduction
The expansion degradation of concrete in this paper refers to the alkali–silica reaction (ASR) anddelayed ettringite formation (DEF). In particular, the ASR is the most common type of concrete
deterioration, with numerous reports at piers worldwide. Both types of degradation are difficult
to control upon occurrence. Artificial lightweight aggregate (ALA) has recently attracted
attention as an ASR countermeasure. However, the mechanism of ASR reduction by ALA is170

Bridge Engineering Institute Conference 2023 (BEI-2023)
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qualitatively and quantitatively unclear, and there are no verified cases of the inhibitory effect of
ALA on DEF. Therefore, the aim of this study was to clarify the mechanism of ASR reduction
by ALA, and to examine the applicability of ALA to DEF, which is a similar type of degradation.
2. Methods and Materials
There are two possible mechanisms by which ALA suppresses ASR: physical and chemical
effects. For the physical effects, alkali silica gel (ASG) may be stored in the coarse voids inside
the ALA to mitigate and detoxify the water absorption and expansion of ASG, and the ALA
crushing may mitigate the expansion pressure of ASR due to its low strength. With respect tochemical effects, it was highlighted that the water contained in the ALA may be released into thepores of the concrete, thereby diluting the alkaline concentration of the pore water, and the ALA
may react with the alkali in the concrete, thereby reducing the alkaline concentration. Moreover,there is no information on whether or not ALA reduces the expansion caused by DEF.
Furthermore, there is a lack of information on the aggregate sizes and amounts for which ALA iseffective in controlling the respective expansions. Therefore, in this experiment, fine and coarse
aggregates of ALA were mixed with the specimens that produced ASR and DEF, respectively.
2.1 Expansion and Mass Measurement
To evaluate the difference in the ASR and DEF reduction effects with respect to different
particle sizes and amounts of ALA, a portion of the aggregate in the concrete of the formulationthat promotes ASR and DEF was replaced with ALA, and the expansion was measured. The
mixture proportions of the specimens are shown in Tables 1 and 2.
For the ASR specimens, the reactive aggregate percentage was fixed (30% in the coarse
aggregate), and a portion of the non-reactive aggregate was replaced with ALA. Given that DEF
Table 1. Mixture proportions of ASR specimen

Chemical Additives

Unit weight (kg/m3)

Water

AS-5
AS-10
AS-20
AS-50
AG-5

160

Cement320

AG-10

Fine Aggregate
Non-reactive

ALA

771
736
667
460

24.2
48.4
96.8
242

Reactive

Non-reactive

ALA

706

0

671

22.3

635

44.6

565

89.3

353

223

309

0

805

AG-20

Course Aggregate

AG-50

Plasticizer
C ×%

AE Agent
C ×%

1.8

0.1

Table 2. Mixture proportions of ASR specimen
Unit weight (kg/m3)
Water

CementFine Aggregate
Non-reactive

ALA

DS-50

485

254

DS-100123

508

Chemical Additives
Course Aggregate
Non-reactive

ALA

1059

0

DG-5

1022

23.4

DG-10

985

46.9

DG-20
DG-50

160

320
847

0

911

93.8

688

235

DG-100318

469

Control

1059

0

171

PlasticizerC×%

AE Agent
C×%

1.8

0.1

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Expansion Ratio (x1000-6)

Expansion Ratio (x1000-6)

Expansion Ratio (x1000-6)

Expansion Ratio (x1000-6)

requires a high temperature
4000
4000
AS5
AS10
in the early stage of aging,
3000
AS20
3000
AS50
the specimens were pre2000
2000
cured in a thermostatic
1000
1000
chamber at 20 °C for 4 h,
AG5
AG10
AG20
and then cured at a high00AG50
temperature of 90 °C for 12
0
10
20
30
40
50
0
10
20
30
40
50
Accelerated Age (week)
Accelerated Age (week)h.
(a) with ALFA(b) with ALCA
2.2 Chemical Method andFig. 1. ASR-induced expansion with ALA
Aggregate Reactivity Test
10000
10000
A chemical method (JIS A
DS50
DG5
DS100
DG10
8000
8000
DG20
1145) was used to confirm
DG50
6000
6000
the reactivity of the reactive
4000
4000
aggregate and ALA. In
2000
2000
particular,
100
gof
aggregate was immersed in
0
0
0
10
20
30
40
50
0
10
20
30
40
50
400 mL of 1 mol/L NaOH
Accelerated Age (week)
Accelerated Age (week)(a) with ALFA(b) with ALCA
solution, and 10 mL of theFig.
2.
DEF-induced
expansion
with ALA
solution was periodicallydrained off to confirm theeffect of the aggregate particle size on thereactivity. The alkali concentration reduction(Rc) was measured by neutralization titration,and the silica elution (Sc) was measured byatomic absorption spectrophotometry.
3. Results and Discussion
AG series
AS series
Figs. 1 and 2 present the results of the expansion
ratio. In the ASR shown in Fig. 1, both thecoarse and fine ALA exhibited a greater decreasein expansion with an increase in the ALA ratio.
Similarly, the DEF results shown in Fig. 2
indicate that ALFA was effective in suppressing
expansion due to DEF, and that the replacementof all fine aggregate with ALA suppressed DEF
DS series
DG series
expansion almost completely. Moreover, unlike
Fig. 3. Results of X-ray CT
ASR, ALCA was not effective in suppressing
DEF expansion.
In this study, X-ray CT imaging was performed on the ASR and DEF specimens after 52 weeks
of acceleration to confirm the internal properties of the concrete. Examples of the results areshown in Fig. 3. The X-ray CT images of the AG series reveal that the entrapped void increased
with an increase in the amount of ALA (white box in left Fig.), whereas the porosity of the AS
series decreased with an increase in the amount of ALA. Moreover, X-ray CT imaging revealed
multiple cracks propagating through the reactive aggregate (red box in right Fig.). However, nocracks penetrated the ALA. Under the conditions employed in this study, the probability of thepremature release of expansion pressure due to the crushing of the ALA was considered as low.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
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4000
Expansion Ratio (x1000-6)

As for the DEF specimens, the specimens mixed
with ALFA showed almost no cracking at all,whereas the specimens mixed with ALCA showed
gaps specific to DEF, not only around the regularaggregate but also around the ALCA.

3000

AS-expAS-simAG-expAG-sim

2000

4. Discussion from Aggregate Reactivity Tests
1000
and ASR Expansion Prediction Model
From the results of the aggregate reactivity test,
0
0
10
20
30
40
50
60
the reactive aggregate and ALA used in thisReplacement rate of ALA aggregate (%)
experiment resulted in a very large Sc in
Fig. 4. Comparison of ASR predictions
comparison with the Rc. Moreover, the alkali
with experimental results.
consumption of ALA increased with a decrease in
the grain size.
In this study, based on the results obtained from the aggregate reactivity test, the ASR expansion
was predicted using numerical model proposed by Uomoto et al. A comparison of predicted andexperimental ASR expansion at week 52 of the accelerated case is shown in Fig. 4. As can beseen from the figure, the numerical predictions did not reproduce the differences between ALFA
and ALCA, although they show a tendency to control the amount of expansion as the ALA
replacement rate increases. In particular, it was difficult to reproduce the effect of ALFA on the
suppression of ASR expansion. Possible reasons why the analysis fails to reproduce the ALFA
trend are as follows. Previous studies confirmed that ALA milled powder undergoes a pozzolanic
reaction. On the other hand, aggregate reactivity tests were performed by immersion in NaOH
solution, and the calcium required for the pozzolanic reaction was not present. Therefore, theconcrete with ALFA underwent a pozzolanic reaction with calcium supplied by the cement, andthe increased alkali consumption prevented ASG due to the reactive aggregate, thus resulting in
the prevention of ASR expansion. Assuming that a similar reaction is occurring in the DEF
specimen, the DEF expansion might be decreased due to a decrease in pH, consumption of
Ca(OH)2, and an increase in mass transfer resistance in the early stages of material aging via the
pozzolanic reaction.
4. Conclusions
In this study, the mechanism by which ALA minimizes ASR and DEF expansion wasinvestigated, and the findings can be summarized as follows:
1) When ALA was used to replace a portion of the concrete in the mixture proportion that
produced ASR and DEF, both ALCA and ALFA reduced ASR expansion. Moreover, only the
specimens with ALFA replacement exhibited decreased DEF.
2) The results of the ASR expansion prediction model indicated that the mechanism of ASR
reduction by ALA is significantly influenced by the consumption of alkali, and that ALFA
underwent a pozzolanic reaction, which resulted in a greater reduction in ASR expansion.
3) The DEF reduction effect was observed only for ALFA, thus suggesting that the pH decreaseat the early stage of the material aging due to the pozzolanic reaction and the consumption of
Ca(OH)2 had a significant effect on the DEF reduction.
5. Acknowledgments
This work was supported by JST FOREST Program, Grant Number JPMJFR215N

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Repair and FRP Composites

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Electromechanical Response of Smart Repair Material with Carbon Black
Based on Rapid Setting Cement
Tae Uk Kim1, Min Kyoung Kim2, Lai Thanh Tu3, and Dong Joo Kim4,5*
1: Department of Civil and Environmental Engineering, Sejong University, 209, Neungdong-ro,
Gwangjin-gu, Seoul 05006, Republic of Korea; email: rlaxodnr1117@gmail.com
2: Department of Civil and Environmental Engineering, Sejong University, 209, Neungdong-ro,
Gwangjin-gu, Seoul 05006, Republic of Korea; email: mkkim9112@naver.com
3: Department of Civil and Environmental Engineering, Sejong University, 209, Neungdong-ro,
Gwangjin-gu, Seoul 05006, Republic of Korea; email: thanhtuxdgt@gmail.com
4: Department of Civil and Environmental Engineering, Sejong University, 209, Neungdong-ro,
Gwangjin-gu, Seoul 05006, Republic of Korea; email: djkim75@sejong.ac.kr
5: Railway Infrastructure Research Center, Sejong University, 209, Neungdong-ro, Gwangjin-gu,
Seoul 05006, Republic of Korea.
*: corresponding author
Keywords: Self-sensing concrete, repair material, rapid setting cement, carbon black
Abstract: This research aimed to develop smart repair materials (SRMs) with self-sensing
capacity at early age (24 hours) by using rapid setting cement and carbon black (CB). CB was
reinforced by 0.5, 1.0, 2.0, and 3.0 wt.% of cement. The electrical resistance of SRMs wasmeasured using the 2-probe method based on alternative current. The self-sensing capacity of
SRMs was highest at 0.5 wt.% CB (CB05). The stress sensitivity coefficient (SSC) of CB05 at
24 hours was measured as 0.373 %/MPa: SSC decreased under more than 0.5 wt.% CB owing to
the stabilization of conductive network. The electromechanical response of CB05 was further
investigated at the age of 3 and 7 hours to investigate the effect of age on the self-sensing
capacity of SRMs. The self-sensing capacity of SRMs was also affected by age owing tocontinued hydration. The SSC of CB05 increased from 0.229 to 0.672 %/MPa (193%) as the age
increased from 3 to 7 hours, while it decreased from 0.672 to 0.373 %/MPa (44.5%) as the age
further increased to 24 hours.
1. Introduction
Many concrete structures around the world have suddenly collapsed owing to early deteriorationof concrete such as the bridge in Genova, Italy in 2018 and the building in Karachi, Pakistan in
2020. One of main reasons for catastrophic failure of concrete structures is concrete crack.
Concrete cracks have to be timely repaired and maintained with suitable methods. Currently,
epoxy resins (Kan et al. 2021) and cement-based materials (Assaad, 2018) have been used to
repair concrete cracks. However, epoxy resin has shown low fire resistance, been poorly cured at
low temperature, and generated cracks at the interface owing to different coefficient of thermal
expansion with concrete structures. Besides, additional sensors or equipment are required toevaluate the quality of repair by using epoxy resin. Although visual inspection, image processing(Choi et al. 2019), and ultra-sonic wave (Kim et al. 2020) have been developed and applied to175Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

inspect the condition of repaired cracks, the reliability of aforementioned method is still quite
low.
This study proposed to develop and apply smart repair materials (SRMs) with selfsensing and rapid setting capacity. The self-sensing capacity of SRMs is expected to evaluate thequality of repaired cracks based on the measured electrical resistance of it in early age (in 24
hours) while the rapid setting property of SRMs is expected to quickly recover the mechanicalperformance of the deteriorated structures in short time. The SRMs containing carbon black (CB)with small particle size can be applied to concrete cracks with 0.3 mm crack width. This researchaims to develop a smart repair material with high self-sensing capacity. Specific objectives are(1) to investigate the effect of CB content on the electromechanical response of SRMs, and (2) todiscover the effect of age on the electromechanical response.
2. Materials and Test Setup
Rapid setting cement and CB were used in the composition of SRMs. The average particle sizeof CB was 120 nm. To investigate the effect of CB content on the strength and self-sensing
capacity of SRMs, the content of CB as 0.0, 0.5, 1.0, 2.0, and 3.0 wt.% by cement was added tothe test series of P, CB05, CB10, CB20, and CB30, respectively. Water to cement ratio was 0.6.
Polycarboxylate based superplasticizer containing 25% solids and 75% water was used toenhance the workability of SRMs.
Figure 1 shows the test setup for the electromechanical response of SRMs. The loading
speed and data frequency were 1 mm/min and 60 Hz, respectively. The electrical resistance of
SRMs was measured by using alternative current (AC) based SI1260 impedance/gain-phaseanalyzer machine (Solatron, 1260 A). The electromechanical response of SRMs was investigatedat the age of 24 hours and that of CB05 was further investigated at the age of 3 and 7 hours.

Fig. 1. Test setup for investigation of the electromechanical response of SRMs
3. Test Results
Figure 2 shows the compressive stress versus fractional change in resistivity (FCR) curves of
SRMs corresponding to different ages, while Table 1 lists the electromechanical parameters of
SRMs. As shown in Fig. 2, the fractional change in electrical resistivity (FCR) of SRMs
generally increased as the compressive stress increased. The electrical resistivity (ρ), FCR andstress sensitivity coefficient (SSC) can be calculated using measured electrical resistance (R) of
SRMs as provided in Eqs. (1), (2), and (3). In addition, the value of SCC was different
corresponding to the different age of SRM.
ρ = R×A/L
FCR = (ρi-ρt)/ρi×100 (%)
SSC (%/MPa) = FCR/σ

(1)
(2)
(3)

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where A is the cross-sectional area (= 2,500 mm3), L is the distance between the two electrodes(gauge length = 25 mm), ρi is the initial electrical resistivity, ρt is the electrical resistivity at timet, and σ is the compressive stress.

Fig. 2. Electromechanical response of CB05 corresponding to different age
Table 1. Electromechanical parameters of SRMs corresponding to different CB contents and age

Notationσ (MPa)ρi (kΩ-cm)
ρc (kΩ-cm) Δρ (kΩ-cm)
FCR (%)SSC (%/MPa)
P
24 h
28.69
22.79
20.46
2.34
10.25
0.354
3h
19.55
6.44
6.15
0.29
4.48
0.229
CB05
7h
23.65
15.16
12.75
2.41
15.89
0.672
24 h
29.79
23.25
20.67
2.58
11.08
0.373
CB10 24 h
27.55
15.50
14.23
1.27
8.21
0.298
CB20 24 h
25.34
17.20
15.82
1.38
8.02
0.316
CB30 24 h
24.22
7.85
7.59
0.27
3.40
0.140
σ: Compressive strength, ρi: Initial electrical resistivity, ρc: Electrical resistivity at peak stress,
Δρ: Change in the electrical resistivity

4. Discussion
4.1 The effect of functional filler on the self-sensing capacity of SRMs
CB05 among the investigated matrices in Table 1 produced the highest SSC (0.373 %/MPa) at
the age of 24 hours. The SSC of SRMs at the age of 24 hours increased from 0.354 to
0.373 %/MPa (5.4%) as the content of CB increased from 0 to 0.5 wt.%, and then it decreased
from 0.373 to 0.140 %/MPa (62.5%) as the content from CB further increased to 3.0 wt.%. With
small content of CB, the SSC increased owing to more number of partially conductive path in the
SRM matrix containing CB. Based on the increased number of partially conductive path within
SRMs, the electrically conductive network within SRMs could be easily changed under externalload. However, partially conductive paths were transformed to continuously conductive paths
with more than 0.5 wt.% CB. The conductive network containing the increased number ofcontinuously conductive paths became hard to be changed under external load, consequently the
SSC decreased.

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4.2 The effect of age on the self-sensing capacity of SRMs
The SSC of CB05 increased from 0.229 to 0.672 %/MPa (193%) as the age increased from 3 to 7
hours, while decreased from 0.672 to 0.373 %/MPa (44.5%) as the age further increased to 24
hours. At the age of 3 hours, SSC was low because the electrically conductive network of SRMs
was stable owing to a lot of water not bound in hydrates. However, the SSC increased because
the electrically conductive network became more easily changed under external load at the age of
7 hours owing to reduction of water with hydration and transformation of the continuously
conductive paths into the partially conductive paths. At the age of 24 hours, and the compressive
strength increased and the number of partially conductive path decreased owing to improvementof density and reduction of water in pores with continued hydration, thus SSC decreased.
5. Conclusions
This study investigated the self-sensing characteristics of the proposed smart repair materials(SRMs) corresponding to different CB contents (0, 0.5, 1.0, 2.0, and 3.0 wt.% by cement) at theearly age (3, 7 and 24 hours). The reinforcement ratio of CB and the age of SRMs clearlyaffected the electromechanical response of SRMs. The following conclusions can be drawn:
 The SRM (CB05) containing CB (0.5 wt.% by cement) produced the highest stresssensitivity coefficient (SSC), 0.373 %/MPa, at the age of 24 hours.
 The amount of CB clearly affected on the SSC of SRM by changing the number of partiallyand continuously conductive electrical paths in the matrix.
 As the age increased from 3 to 7 hours, the SSC of CB05 increased from 0.229 to
0.672 %/MPa (193%) because of reduction in the water of matrix and the transformation of
the continuously conductive paths to the partially conductive paths. However, it decreased
from 0.672 to 0.373 %/MPa (44.5%) as the age further increased to 24 hours owing to thecontinued hydration.
Further research about the quality of crack repair and the self-sensing capacity of SRMs isrequired under different environments.
6. Acknowledgments
This work is supported by the Korea Agency for Infrastructure Technology Advancement(KAIA) grant funded by the Ministry of Land, Infrastructure and Transport (Grant RS-202200142566).
7. References
Assaad J.J. 2018. Development and use of polymer-modified cement for adhesive and repair
applications, Construction and Building Materials, 163, 139–148.
Choi, Y.S., Kim, J.H., Cho, H.C., Lee, C.J. 2019. Asphalt concrete pavement surface crack
detection using convolutional neural network, Journal of the Korea Institute for Structural
Maintenance and Inspection, 23(6), 38-44.
Kan, Y., Lee, M., and Lee, H. 2021. Experimental investigation of mode-I fracture toughness ofreal-cracked concrete repaired by epoxy, Construction and Building Materials, 293, 123490.
Kim, J.H., Hong, J.Y., Kim, R.R., Woo, U.W., Choi, H.J. (2020) Development and applicationof IoT-based contactless ultraosonic system. Journal of the Korea Institute for Structural
Maintenance and Inspection, 24(3), 70-79.

178Bridge Engineering Institute Conference 2023 (BEI-2023)
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Bending Tests of GFRP Sandwich Panel Using a Rigid Polyethylene Core
Hiroki Iwata1, Sumitaka Inoue2, and Masahide Matsumura3*
1: Kumamoto University, Kumamoto, Japan; email: 223d8301@st.kumamoto-u.ac.jp: COMTEC Inc., Kumamoto, Japan; email: inoue@compotec.jp
3: Kumamoto University, Kumamoto, Japan; email: matsumura-m@kumamoto-u.ac.jp
2*: corresponding author

Keywords: GFRP; polyethylene; sandwich panel; water absorption; three-point bending tests
Abstract: Sandwich plate panels, consisting of high-strength skin and lightweight core, are widelyused in various fields such as airplanes, automobiles, and ships. Even for deck slabs of foot bridges,a sandwich panel made of high durable GFRP plates for the skins and of balsawood for the core is
generally used for practical use the panel of lighter weight and of higher durability. However,previous studies pointed out that load-bearing capacity of the sandwich plate using the balsawood
for the core might decrease when exposed to rain and snow for a long period, as strength andbending stiffness of water absorbed balsawood take smaller values compared with dried one. Thenfocused on in this study is a GFRP sandwich panel using a rigid polyethylene, having an
independent foamed structure and showing low water absorption, for the core material to develop
a higher durable deck slab for the foot bridges. The GFRP plates and the polyethylene core areintegrated by using the VI molding technique in which resin is impregnated by vacuum pressure.
Water immersion test, and three-point bending tests were conducted to check water absorption
resistance and to understand the load-bearing capacity and the failure mode of the sandwich panels.
The test results revealed that the GFRP sandwich panels show a good water absorption resistanceand the shear failure in the core polyethylene were observed in all the specimens in the loading
test, but a good applicability for structural use as the deck slab of the foot bridges within 1.2m inthe span length.
1. Introduction
The Japanese climate is hot and humid, and wood decay fungi tends to grow in this environment.
So, deterioration due to corrosion of wooden pedestrian slabs have been reported, then needs forslab replacement are increasing, but when pedestrian bridges are located in narrow areas such as
mountainous regions, it is difficult to use heavy machines for construction. Therefore, a new slab
having excellent durability and a lightweight structure transported by hand can be developed. Then,the authors focused on sandwich structures, of lightweight and high rigidity. Sandwich structures
are widely used in various fields such as aircraft, automobiles, and ships. Even for deck slabs ofpedestrian bridges, a sandwich panel made of high durable GFRP plates for the skins andbalsawood for the core is generally used to realize the panel of lighter weight and of higher
durability (Ishii et al. 2019). However, since the strength and modulus of elasticity of balsa
materials decreases due to water absorption, the load-bearing capacity of balsa plated member may

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2. Water Immersion Test
Two test pieces (100 × 100 × t36 mm) were
submerged into water at 20 ℃ in a thermostaticchamber during for 60 days. The dry mass isdetermined after the samples are dried in a hot-air dryer
at 65 ℃ for three days. A water absorption rate wascalculated from the Eq. (1) by measuring the massusing an electronic scale.
𝑄𝑄 =𝑀𝑀2 −𝑀𝑀1
𝑀𝑀1

× 100

(1)

GFRP plate(6mm)

Rigid polyethylene(25mm)

Fig. 1. Laminated configuration of GFRP
sandwich panel with rigid polyethylene core

(a)

(b)

(c)

(d)

(e)
Fig. 2. Failure modes of sandwich panels
subjected to bending moments (Miyairi 2019)

1.2
Water absorption rate (%)

decrease when exposed directly to rain or snow for along period (R.L. Sadler et al. 2009). Then the load
bearing capacity of the sandwich plate using thebalsawood core also may decrease exposed under hot
and humid environments. Then focused on in this study
to develop a GFRP sandwich deck for the foot bridges
using a rigid polyethylene for the core material, having
an independent foamed structure, and showing a lower
water absorption. Fig. 1 shows the GFRP sandwich
panel with rigid polyethylene core. The thickness of thesurface material GFRP plate as is 6 mm, and thethickness of the core material rigid polyethylene as is
25 mm. The GFRP plates and the polyethylene core areintegrated by using the VI molding technique in whichresin is impregnated by vacuum pressure. When the
sandwich panel is used as deck slabs, then the panel isset between longitudinal members or between
transverse ones bending moment is subjected. Failuremodes of shown in Fig. 2. (a) and (b) are skin panel
failures, (c) and (d) are core failures and (e) isdelamination of bonding surface. In particular, failuremode (e) requires sufficient bond strength is expected
due to water immersion because and arises a sudden
loss of bearing capacity the skin and core materials can.
Then, water immersion test and three-point bending
tests were conducted to check resistance against water
absorption resistance and to understand the failuremode of the sandwich panels.

1
0.8
0.6
0.4

0.2
where Q = water absorption rate; M1 = dry mass; M2 =wet mass.
0
Fig. 3 plots changes of averaged value of the water
0
10
20
30
40
50
60
absorption rate shows. A rapid increase of the water
Water Immersion Time (days)
absorption rate was observed in the first few days, andthe maximum value was obtained 1.03 % after 60 days Fig. 3. Averaged value of water absorption

rate

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L=1,000

90

90

O50
Fig. 5. Loading method of 3-point bending test
(unit in mm)
(a)

(a)

(b)
Fig. 4. Water immersion test GFRP sandwich
panel after: (a) outer surface, (b) cut surface(b)

3. Three-Point Bending Tests
Figure 5 illustrates the loading method of threepoint bending tests. The concentrated load isapplied at the center of the specimen. A
displacement transducer is set under the load point.
Fig. 6 shows the damage at the interface after thetest of the sandwich panels, No.3 and No.4 asexamples. No failure of the GFRP plate and theadhesive between the GFRP plate and the rigid
polyethylene was observed. The brittle shear
failure and delamination in the rigid polyethylene
were observed among most of the specimens,
including No.3 at the maximum load. Only thedelamination occurred of No.4. Fig. 7 pictures theloading point before and after the test, of No.2,
where delamination occurred at the position of theinterwoven glass fiber layer in the rigid
polyethylene near the upper interface. Figure. 8

181

(a)

(b)

Fig. 7. Failure at the loading point before / after
the test
50
40

Load (kN)of the water immersion. Previous studies revealed
that the mass of the balsa wood increased to about
six times after 60 days of water immersion, and thatthe rigid polyethylene, consisting of independent
bubble structure had excellent water-absorption
resistance. Fig. 4 shows the surface observation
outer surface and cut surface of the test piece after
the water immersion test. These demonstrate thatwater penetrating was not observed the inside into
but slightly at the interface.

Fig. 6. Failure mode of sandwich panels after
3-point bending tests: (a) No3, (b) No.4

30
20

No.1
No.2
No.3
No.4

10
0

0

15

30
45
60
Displacement (mm)75

Fig. 8. Load-displacement relationshipBridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

plots the load–displacement relationshipobtained from the experiment. The results are
summarized in Table 1. Small differences in theinitial bending stiffness. The classification of
fracture modes is shown in Fig.2. No difference
in experimental values due to differences infracture mode was observed.
Then, Figure 9 illustrates the positions of thestrain gages attached to the side and top andbottom surfaces. Figure 10 plots the strain
distribution in the direction of the cross-sectional
thickness at 5~25 kN. GFRP is more distorted
than rigid polyethylene, and the assumption ofplane retention is apparently not valid. The larger
the load, the stronger the tendency.

Table 1. Summary of the test resultsNo.
12

Displacemnet
Max.Load Coefficient of
at Max.load(kN)variation
(mm)
39.11
62.21
37.45

3

31.28

4

29.91

Avg.

34.44

0.11

-

(c) + (d)66.11(c) + (d)49.00(c) + (d)

50.85

(d)

57.04

-

(c): shear failure, (d): delamination
25

4.75

6

2.5*6
=15
37 25
5*2=102.5
4.75

6

Fig. 9. Positions of strain gages attached to theside and top and bottom surfaces (unit in mm)
5kN

Distance from thickness center (mm)4. Conclusions
In this study, water immersion tests and threepoint bending tests were conducted check water
absorption resistance and to understand the loadbearing capacity and the failure mode of GFRP
sandwich panel. The following conclusions weredrawn from the results of this study:
From the immersion test, it was confirmed that
GFRP sandwich panels using rigid polyethylene
as the core material have excellent water
resistance.
From the strain distribution in the crosssectional thickness direction, it was confirmedthat the bending stress was sheared by the GFRP
plate and reached its final stage throughdelamination of the glass fibers in the rigid
urethane and the shear failure of the rigid
polyethylene.
No peeling of the adhesive surface between
GFRP and rigid polyethylene occurred, and it
was confirmed that there was sufficient adhesivestrength.

Fracture
mode

16.5

10kN

15kN

20kN

25kN

GFRP

11.5
6.5
1.5

Glass fiber
Rigid polyethylene

-3.5
-8.5
-13.5

Glass fiber
GFRP

-18.5
-9000 -7000 -5000 -3000 -1000 1000

3000

5000

7000

9000

Strain (×10-6 )Fig. 10. Strain distribution in cross- sectional
thickness direction

5. References
Ishii et al. 2019. An experimental study of GFRP-Balsa decks for footbridge slab, Journal of
Structural Engineering, 65A, 786-798. (in Japanese).
R.L. Sadler et al. 2009. Water immersion effect on swelling and compression properties of EcoCore, PVC foam and balsa wood, Composite Structures 90, 330-336.
Miyairi. 2009. Sandwich structure. (in Japanese)

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Analysis of Flexural-Bond Behavior of Beams Reinforced with CFRP
Dongkyu Lim1 and Myoungsung Choi2*12: Dankook University, Yongin, South Korea; email: ldk6482@gmail.com: Dankook University, Yongin, South Korea; email: choims@dankook.ac.kr*: corresponding author

Keywords: CFRP bar; Bond strength; Flexural-bond; BS EN 12269
Abstract: Research and application of carbon fiber reinforced polymer (CFRP) as a substitutefor reinforcement is increasing in the field of reinforced concrete, which has been uniformly
established in the design of existing structures. CFRP reinforcement has superior material
properties such as high tensile strength, lightness, and non-corrosiveness compared to steel, but
excessive deflection and brittle failure due to low elastic modulus and elastic behavior are
suggested as problems. This study conducted the 'Flexural-bond test' proposed by BS EN 1229 toevaluate the bonding characteristics of beams subjected to bending loads as a basic study forevaluating the structural performance of members using CFRP as reinforcement. As a result ofconducting an experiment with steel and CFRP as reinforcement variables, the CFRP
reinforcement member showed higher load resistance to bending load than steel, while the bondstrength tended to be lower. Based on this, we tried to closely analyze factors that affect bondcharacteristics by using concrete strength, embedded length, and diameter of reinforcement asvariables.
1. Introduction
Fiber Reinforced Polymers (FRP) are a composite of fibers with excellent mechanical properties
and a matrix that constrains, molds, and transmits stress. Research and application of carbon
reinforced polymers (CFRP) as reinforced substitute materials is on the rise in the field of
reinforced concrete, which has been uniformly established in the design of existing structures. As
shown in Figure 1, CFRP reinforcement have superior material properties such as high tensile
strength, light weight, and non-corrosiveness compared to rebars, but excessive deflection andbrittle destruction due to low elastic modulus and elastic behavior are suggested as problems.
For this reason, an accurate understanding of material properties is required to be applied to an
actual structure. In order for CFRP to be used as a reinforcement substitute, structural
performance by sufficient bond must be secured during the synthesis behavior with concrete. In
other words, it is judged that quantitative evaluation of bond characteristics is necessary for the
integrated behavior of reinforcement-concrete. Therefore, this study conducted an experimentand FEM analysis to evaluate the bond characteristics of structures under flexural load to verify
the structural performance of CFRP reinforcement, and compared and analyzed the flexural-bondbehavior of rebars and CFRP reinforcement.

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Fig. 1. Stress-strain curve by reinforcement type
2. Experimental Method and Materials
In this study, a beam-shaped test specimen with an arch proposed by BS EN 12269 was
fabricated, and an experiment was conducted by simulating a flexural-bond behavior similar toan actual beam structure. Concrete beams were made with a design compressive strength of 35
MPa, and rebars and CFRP reinforcement were set as experimental variables. The
reinforcements used have the same diameter of D13 (12.7 mm), and the yield strength of thesteel rebar and CFRP bars were confirmed to be 400 MPa and 1850 MPa, respectively. The
details and material properties of the test specimens and measuring gauges used are shown in
Figure 2 and Table 1.

(a)

(b)
Fig. 2. Test specimen and gauge setting detail: (a) Conceptual diagram of test specimen(BSEN12269); (b) Specimen setting view184

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Table 1. Girder details
Concretef ck : 30 MPa, ε cu : 0.0033, Ec : 29,100 MPa

CFRP barf f : 1,800 MPa, E f : 130 GPa, Diameter : D13(13 mm)

Steel barf y : 450 MPa, Es : 210 GPa, Diameter : D13(12.7 mm)

f ck : Design strength of concrete

ε cu : Ultimate strain of concrete

Ec : Elastic modulus of concrete
f : Tensile strength of CFRP barf

E f : Elastic modulus of CFRP bar
f y : Yield strength of steel bar

Es : Elastic modulus of steel bar
3. Result and Analysis
The average experimental results of each test specimen are shown in Table 2, and Figure 3 is agraph showing the load-displacement and load-strain relationship according to the reinforcement
variables. The CFRP reinforcement specimen showed a maximum load of about 145% comparedto the steel specimen, and the deflection amount was about 30%, confirming that brittle failure
occurred after elastic behavior. The ultimate bond strength was defined as the bond stress at thetime of concrete failure, and the CFRP reinforcement specimen was 17.29 MPa and 10.91 MPa,
respectively, which were slightly lower in the rebar and CFRP reinforcement specimens. This isconsidered to have resulted in the yield of rebars and concrete crushing due to the concentrationof stress on the rebar due to the high transverse restraint pressure on the steel when sufficientdevelopment length is secured. In the case of CFRP, splitting occurred in the vertical concrete
cover due to the Wedge and Prying action. However, in order to accurately evaluate the bondstrength, the reinforcement should not yield and pull-out or splitting should occur. Therefore, it
is judged that experiments and analysis that supplement this should be conducted in futurestudies.
Table 2. Experimental result for Flexural-bond test
Ultimate load
No.
Specimen
Failure mode(kN)
1
CFRP(D13)
51.85
SplittingConcrete crushing /2
Steel(D13)35.36
Steel yielding

185

Ultimate bond strength
(MPa)
10.91
17.29

Bridge Engineering Institute Conference 2023 (BEI-2023)
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(a)

(b)

Fig. 3. Flexural-bond test results: (a) Relationship between load and deflection; (b) Relationshipbetween load and strain
4. References
BS EN 12269. 2001. Determination of the bond behaviour between reinforcing steel andautoclaved aerated concrete by the "beam test" Short term test.
Biscaia, H. C., Chastre, C., Borba, I. S., Silva, C., & Cruz, D. 2016. Experimental evaluation ofbonding between CFRP laminates and different structural materials. Journal of Composites forConstruction, 20(3), 04015070.
Hassan, T., & Rizkalla, S. 2004. Bond mechanism of NSM FRP bars for flexural strengtheningof concrete structures. ACI Structural journal, 101(6), 830-839.
Oh, H., Kang, T. S., & Oh, K. C. 2014. A flexural bonding characteristic of GFRP rebar
embedded in concrete beam under cyclic loading. Journal of the Korean Society of Hazard
Mitigation, 14(4), 51-58.
5. Acknowledgements
This work is supported by the Korea Agency for Infrastructure Technology Advancemen(KAIA)
grant funded by the Ministry of Land,Infrastructure and Transport (Grant RS-2021-KA163381).

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Evaluation of Tensile Performance of CFRP Cable Anchorage
Systemusing Non-Corrosive Steel Material
Seung-Hyeon Hwang1*, Ji-Young Kim2, Jin-Young Yoon3,
Tae-Kyun Kim4, and Woo-Tai Jung5
1: Korea Institute of Civil Engineering and Building Technology, Goyang, Gyonggi-do, Republicof Korea; email: shhwang@kict.re.kr
2: Korea Institute of Civil Engineering and Building Technology, Goyang, Gyonggi-do, Republicof Korea; email: jykim0421@kict.re.kr
3: Korea Institute of Civil Engineering and Building Technology, Goyang, Gyonggi-do, Republicof Korea; email: jyyoon@kict.re.kr
4: Korea Institute of Civil Engineering and Building Technology, Goyang, Gyonggi-do, Republicof Korea; email: kimtaekyun@kict.re.kr
5: Korea Institute of Civil Engineering and Building Technology, Goyang, Gyonggi-do, Republicof Korea; email: woody@kict.re.kr*: corresponding author

Keywords: carbon fiber-reinforced polymer; compression sleeve; non-corrosive steel material;anchorage system
Abstract: The anchorages used to fix and connect the cables to structures typically use steel,making it essential to minimize corrosion. The study proposes a fully corrosion-free cable systemthat includes anchorages, as carbon fiber-reinforced polymer (CFRP) cables, which have hightensile strength and are not affected by corrosion. The study modified the details of the sleeveanchorage proposed by Kim et al., which used SM45C steel material, to use duplex steel material,
which has no effect on corrosion. The proposed CFRP cable sleeve anchorage using noncorrosive steel material was evaluated using tensile performance tests to determine its slip in theanchorage, failure mode, and stress-strain relationship. The results showed that the proposed
CFRP cable sleeve anchorage using non-corrosive steel material performed almost as well as the
sleeve anchorage of SM45C steel material, demonstrating that the new system can be used as afully corrosion-free cable system that includes anchorages.
1. Introduction
The anchorage, which is generally a conventional cable and is crimped or swaged, is difficult toensure complete corrosion resistance as it is usually used with plating or coating after machining.
However, if the internal corrosion resistance of the anchorage can be ensured, a cable system thatis not affected by corrosion can be achieved with carbon fiber-reinforced polymer(CFRP) cable,
which is not affected by corrosion. Duplex, a type of stainless steel, contains a large amount ofstrengthening elements such as chromium, molybdenum, and nickel, and is a type of steel withexcellent resistance to chloride corrosion, SSC, pitting, crevice corrosion, wear, and erosion(Rabiet al., 2022). Additionally, this type of steel has excellent mechanical properties because the sizeof the structure is smaller than that of other steel materials. In this study, an anti-corrosive steel

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will be used to find the optimal specifications for a carbon fiber cable anchorage and to suggest acompletely non-corrosive cable system accordingly.
2. Experimental Investigation
The non-corrosive anchorage for CFRP cables refers to the anchorage specifications of SM45C
steel proposed by Kim and Jung (2022). The total length of the anchorage is 135.0 mm, with acentral section at 95.0 mm and the points where fine angles are given on both sides being 20.0mm each. The outer diameter of central section is 25.9 mm, the outer diameter at the end point is
24.0 mm, and the inner diameter is symmetrical at 11.0 mm on both sides. The experimental
variables for the development of non-corrosive anchoring devices for CFRP cables were set atintervals of 0.1 mm for the outer diameter( do ) of central section in the range of 25.5 - 25.9 mm.
CFRP cables were manufactured in lengths of approximately 1,000 mm, with three samples foreach variable, and the anchorages were installed at positions about 100 mm away from both endsof the CFRP cable. An aluminum insert was set between the CFRP cable and the anchorage.
Tensile tests of the CFRP cable system were conducted using a 1,000 kN capacity UTM, with adisplacement control rate of 5 mm/min.
3. Test Results and Discussion
The failure mode of the specimen, which had a do of 25.5 mm for the anchorage, resulted inslip between the anchorage and the CFRP cable, leading to the termination of the experiment(Fig.
2(a)). The failure mode of the specimens, which had an do of 25.6 mm and 25.7 mm for theanchorage, mostly showed a flower-shaped fracture(Fig. 2(b) and (c)). The failure mode of thetest specimens, which had an do of 25.8 mm and 25.9 mm for the anchorage, showed a fracture
at the inner end of the single anchorage system between the anchorage and the CFRP cable(Fig.
2(d) and (e)).

(a) do = 25.5 mm(b) do = 25.6 mm(c) do = 25.7 mm(d) do = 25.8 mm(e) do = 25.9 mm

Fig. 1. Failure mode of CFRP cable anchorage systems
The tensile test results according to the variable outer diameter of the anchorage are shown in
Table 1. The applied load-displacement relationship of the specimen, which had an do of 25.5mm for the anchorage, showed linear behavior up to a displacement of approximately 15 mm,
and thereafter, the displacement increased due to slip, but the load remained constant(Fig. 2).

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The applied load-displacement relationship of the specimen, which had an do of 25.6 mm for theanchorage, was similar to that of the specimen with an anchorage do of 25.5 mm, but the amountof slip decreased. However, the applied load-displacement relationship of the specimens withanchorage do of 25.7 mm, 25.8 mm, and 25.9 mm increased linearly up to the maximum load.
The axial stress-strain relationship of all specimens increased linearly up to the point of reaching
the tensile strength(Fig. 3). The average tensile strength of the specimens with do of 25.6 mm
and 25.7 mm was the highest at approximately 3,143 MPa and 3,140 MPa, respectively, within
the do range of 25.5 - 25.9 mm(Fig. 4).
Table 1. Test results
Specimen
Outer diameter
25.5 mm
Outer diameter
25.6 mm
Outer diameter
25.7 mm
Outer diameter
25.8 mm

300

3500

250

3000

Elastic modulus(MPa)
182,892
181,810183,021
182,574
184,917
183,540
182,202
183,553
185,191
181,723
183,111
183,342
183,082
186,259
182,155
183,832
174,037
170,681
172,250
172,323

2500

200

peak

100
50
0

5

10
15
20
diaplacement (mm)

25

peak

2000

do = 25.5 mm
25.5
do = 25.6 mm
25.6
do = 25.7 mm
25.7
do = 25.8 mm
25.8
do = 25.9 mm
25.9

150

0

Strain(µ )
14,375
14,782
15,022
14,726
16,82517,024
17,53017,12617,141
18,082
16,16817,13015,12915,441
16,54315,704
15,05816,037
15,101
15,399Tensile strength
(MPa)
2,629
2,688
2,749
2,689
3,111
3,125
3,194
3,143
3,174
3,286
2,961
3,140
2,770
2,876
3,013
2,886
2,621
2,737
2,601
2,653

Axial stress (MPa)Applied load (kN)

Outer diameter
25.9 mm#1#2#3
Average#1#2#3
Average#1#2#3
Average#1#2#3
Average#1#2#3
Average

Applied load
(kN)
206.5
211.1
215.9
211.2
244.4
245.4
250.9
246.9
249.3
258.1
232.5
246.6
217.5
225.9
236.7
226.7
205.8
215.0
204.3
208.4

1500
1000

30

Fig. 2. Applied load-displacement relationship

189

500
0

0

5000

10000
Strain (×10-6)

do = 25.5 mm
25.5
do = 25.6 mm
25.6
do = 25.7 mm
25.7
do = 25.8 mm
25.8
do = 25.9 mm
25.9
15000

20000

Fig. 3. Axial stress-strain relationshipBridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023
3500

Axial stress (MPa)

3000
2500
2000
1500
1000
500
0

25.5

25.6

25.7
do (mm)

25.8

25.9

Fig. 4. Tensile strength results according to outer diameter
The average tensile strain of the specimens with do of 25.6 mm and 25.7 mm was also thehighest at approximately 17,126 µ and 17,130 µ , respectively, within the same range. The
average tensile strain of the specimens with do of 25.5 mm, 25.8 mm, and 25.9 mm wasapproximately 14,726 µ , 15,704 µ , and 15,399 µ , respectively, and was proportional to thedecrease in tensile strength of the specimens with do of 25.6 mm and 25.7 mm. The average
elastic modulus of the specimens within the do range of 25.5 - 25.8 mm was approximately
183,000 MPa.
4. Conclusions
Based on tensile tests on the outer diameter of the CFRP cable non-corrosion anchorage system,
the following conclusions were inferred:
1. The failure mode of the specimens with an outer diameter ranging from 25.6 - 25.7 mm
for the anchorage system mostly showed a flower-shaped fracture.
2. The applied load-displacement relationship of the specimens with an outer diameter of
25.6 mm or less for the anchorage system showed slip between the CFRP cable and theanchorage, while the specimens with an outer diameter of 25.7 mm or greater showed alinear increase up to the maximum load.
3. The average tensile strength of the specimens with an outer diameter of 25.6 mm and25.7 mm for the anchorage system was the highest, 3,143 MPa and 3,140 MPa,
respectively.
4. The optimal outer diameter for the non-corrosion anchorage system for CFRP cables is25.7 mm.
5. Acknowledgments
Research for this paper was conducted under the KICT Research Program (project no. 20230110-001,
Development of high performance construction materials based on carbon (focus on insulationmaterials for building and corrosion free cable)) funded by the Ministry of Science and ICT.
6. References
Kim, T.K. and Jung, W.T. 2022. Improvement of Anchorage Performance of Carbon FiberReinforced Polymer Cables, Polymers, 14(6), 1-18.
Ravi, M., Shamass, R., and Cashell, K.A. 2022. Structural Performance of Stainless Steel
Reinforced Concrete Members: A Review, Construction and Building Materials, 325(28), 1-14.

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Failure Mode and Adhesive Strength of Single Lap Joint of GFRP LaminatedPlates with Different Fiver Volumes Bonded by a Low Elastic Adhesive
Masahide Matsumura1*, Sumitaka Inoue2, Hiroki Iwata3, and Yuto Anan3
1: Kumamoto University; Kumamoto Japan, email: matsumura-m@kumamoto-u.ac.jp: COMTEC Inc., Kumamoto Japan; email: inoue@compotec.jp
3: Kumamoto University; Kumamoto Japan, email: 223d8301@st.kumamoto-u.ac.jp
4: Kumamoto University; Kumamoto Japan, email: 219d8302@st.kumamoto-u.ac.jp*: corresponding author
2

Keywords: adhesive strength; single lap joint; GFRP; fiver volume; low elastic adhesive
Abstract: In this study, tensile tests of single lap joints, where two of GFRP laminated plates withdifferent fiber volume are bonded by a low elastic adhesive, are carried out to discuss influencesof fiber volume differences in the loading direction and of elastic modulus of the adhesives on
failure mode and on adhesive strength of the joints. The GFPR plates used in the test consist of 13
and 22 layers having almost the same fiber volume in total; fiberglass mat of 1 direction and glass
roving cloth are alternately laminated. Then test results indicated that failure mode changes
depending on the fiber volume in the loading direction and the adhesive strength takes smaller
value and fluctuates when the fiber volume takes smaller and the failure mode is delamination of
the layers.
1. Introduction
Developments of a reliable joint connecting FRP members are keys to develop a longer span of
FRP bridge structure. The adhesive strength is to be experimentally evaluated by tensile lap-shear
test program using of the single lap specimen illustrated in Fig. 1. However, a lot of parameters
influence on the test results such as failure modes, adhesive strength of single lap joint of GFRPs.
This is because a bending moment due to an eccentric axial load, causing a rotation of theconnected part, is subjected to the adhesive connection between the GFRP plates, and flexural andaxial stiffness of the plates greatly influence on these tested values.
Then in this paper, GFRP plates molded by the Vacuum Infusion technique at room
temperature having different fiber volumes, consisting of 13 and 22 layers of fiberglass mats andglass roving cloths, and a low elastic adhesive of 1 mm in the thickness are prepared for the test.
The fiber volumes are varied by using 3 of different GF sheets commercially provided and 2 ofthe directions of 0 and 90 degrees.

(a)

(b)

Fig. 1. Lap-shear test (a)test specimen, (b) expected deformation

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2. Tensile Lap-Shear Test
The adhesive strength is obtained as the maximum load divided by the adhesive area, by the single
lap-shear test. Here, the fiber volumes are varied by using 3 different fiber types; A to C of GF
sheets commercially provided as listed in Table 1. As listed in Table 2, tensile modulus of the glass
fibers take different values from 40 to 76, based on nominal values provided from suppliers. GFRP
plates were alternate laminated structure of glass mats oriented in 1 direction and of roving clothin 4 directions, and 13 layers for specimens A and B and 22 layers for specimen C because of thefiber volume per one sheet of Fiber type C is smaller. Tensile test results of the GFRP Plates of 3
fiber types are summarized in Table 4. Vinyl ester resin in Table 5 is used as adhesive and the
GFRP plates were molded by the VI molding technique.
Test specimens illustrated in Fig. 2 are prepared. Both the bonding length and the width
were 100 mm respectively, and a low modulus adhesive with an elastic modulus of 170 to 345
MPa, as listed in Table 7, mentioned later, was used for the test. The target adhesive thickness was
1 mm controlled by using glass beads of 1 mm diameter. Strain gauges were installed at thepositions also shown in Fig. 2. 1,000 kN universal testing machine was used for the tests, and theloading speed was manually adjusted to 6 mm/min by reference to JIS K6850.
Table 1. Fiber types and test cases
Cases
A-0-No.
A-90-No.
B-0-No.
B-90-No.
C-0-No.
C-90-No.
a

Fiber type
A
A
B
B
C
C

Fiber angle
0°
90°
0°
90°
0°
90°

Numbers

Table
Fiber
types and properties
Table 2.2.Fiber
types

a

σt
(GPa)
1.9-2.4
SELCOM UNIASELCOM QI
1.9-2.6
JA EUL
0.9
B
JA EQX
0.9
SHINDO W600GCSHINDO F880G
σt = Tensile strength
E t = Tensile modulus
Fiber Type5

: Numbers of specimen

Table 4. Adhesive properties

Table
3. Fiber volumes and fiber directions
Table 3. Fiber volumes and fiber directions
Dir.
Fiber typeA
B
C

Table 5. Adhesive properties
σt
Et
τ
(GPa)
(GPa)
(MPa)
MA560-1 17.2-20.7 172-345 11.7-15.2
σt = Tensile strength
E t = Tensile modulus
τ = Lap-shear strength
φ = Elongnation ratio

total
2(g/m )
9300 2100 2100 2100 15600
8928 2107 2100 2407 15542
9764 2180 2030 2708 16682
0°

+45°

-45°

90°

F
1

2 3

Et
(GPa)
69-76
73
40
40
70
70

5

7 8 9

B

Fig. 2. Shape of test specimen (unit: mm)
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φ
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3. Experimental Results
Table 5 shows the fracture modes of the specimens; an adhesive failure, AF, where the adhesive
adhered to both surfaces of the GFRP plates like specimen A-0-No.5 and a GFRP delamination
failure, DF, where the GFRP plate of one side delaminates like specimen A-90-No.2.
Table 6 summaries material test results of the GFRP plates and lap-shear test results by using the
averaged values. Fig. 3 plots the relationship between the maximum loads and fiber volumes, with
Table 5. Failure modes
0

Failure Mode

90

A-0-No.5

A-90-No.2

AF

DF

A

B-0-No.5

B-90-No.4

DF

DF

B

C-0-No.5

C-90-No.1

AF

DF

C

150

Table 6. Test results
(GPa)

(GPa)

(g/m )

A-0

558

30.2

9300

A-90

230

17.6

2100

B-0

570

32.5

8928

B-90

191

14.7

2407

C-0

513

30.8

9764

C-90

220

17.9

2708

2

a: Fiber volume of loading direction
b: thickness of adhesive layert (mm)
1.076
1.086
1.098
1.309
0.913
1.224
1.766
1.511

(kN)
116.4
90.0
105.1
101.1
71.7
117.8
107.1
102.7

(MPa)
11.6
9.0
10.6
10.0
7.1
11.7
10.5
10.2

125

Failure
Mode
AF
DF
AF
DF
AF
DF
AF
DF
AF
DF
AF
DF

100

P max (kN)

Table 4. Material test results and Tensile share test results
τmax
Et
σt
Cases
Fibera thicknessb P max

75
50

A-0_AF
A-90_DF
B-0_AF
B-0_DF
B-90_DF
C-0_AF
C-90_AF

25
0

0

2000

4000

6000

AW

(g/m2)

A-0-avg(AF)A-90-avg(DF)B-0-avg(AF)B-0-avg(DF)B-90-avg(DF)C-0-avg(AF)C-90-avg(AF)8000

Fig. 3. Pmax – Aw relationship

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the fracture modes of each specimen. The figure also plots the averaged value of the maximum
load among which the same fracture modes was observed.
These test results explain that the maximum load becomes larger as the fiber volume
increases, regardless of the differences of the fracture modes when the same GFRP fibers are used
but the loading directions are different. When compared with the dame GFRP fiber specimens, themaximum loads with AF take larger than those with DF. The specimen B-0, which takes the lowest
elastic modulus and fiber volume in the 0° direction as indicated in Tables 2 and 3, 4 of specimens
B-0 showed DF and the averaged strength of B-0 take the smallest among all the cases, while thefracture modes of the specimens A-0 and C-0 were AF. On the other hand, the maximum load of
B-90, taking a larger fiber volume than A-90, was smaller than that of A-90.
Figure 4 shows bending moment diagrams at the maximum load, calculated from measured
strains on the front and back surfaces. The figure indicates that the failure starts from the bondingedges where the bending moments become larger and the maximum bending moment was smaller
among the specimen with DF of the failure mode.

A-0
A-90
B-0
B-90
C-0
C-90

600000M (N・mm)

4000002000000-200000-400000-600000

-50

0

50

100

150

X (mm)

200

250

300

Fig. 4. Bending moment diagram at Pmax
4. Summary
Tensile shear tests of GFRP single-lap adhesive bonded joints were conducted in this study byfocusing on fiber volumes of glass fibers. The following is a summary of the results.
The test results indicate that the bonding strength and fracture modes vary depending on themechanical properties of the glass fiber and the fiber volumes of the GFRP plates. Two types offracture modes; Adhesive Failure, AF and Delamination Failure, DF, were observed in the tests.
The maximum loads of AF tend to be higher as the fiber volumes of the GFRP plate take larger.
On the other hand, DF was observed when bending moment subjected to the joints is smaller, forinstance, when the elastic modulus of the GFRP plates take smaller. The bonding strength andfracture modes were different depending on the physical properties of the fibers and the fiber
volume.

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Ultra-High Performance ConcreteBridge Engineering Institute Conference 2023 (BEI-2023)
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Titanium Alloy Reinforced Ultra-High Performance Concrete (TARUHPC):A Next Generation of Advanced Materials for Civil Infrastructure
Mahesh Acharya1, Jared Cantrell2, Luis Bedriñana3, and Mustafa Mashal4*
1: Idaho State University, Pocatello, Idaho, USA; email: achamahe@isu.edu: Idaho State University, Pocatello, Idaho, USA; email: cantjare@isu.edu
3: Universidad de Ingenieria y Tecnologia – UTEC, Lima, Peru; email: lbedrinana@utec.edu.pe
4: Idaho State University, Pocatello, Idaho, USA; email: mashmust@isu.edu
2*: corresponding author

Keywords: TARUHPC; ensemble learning; machine learning; UHPC; pull-out test; bondstrength prediction.
Abstract: The paper introduces a next generation of composite advanced materials titled“Titanium Alloy Reinforced Ultra-High Performance Concrete” (TARUHPC). In this solution,
Titanium Alloy Bars (TiABs) and Ultra-High Performance Concrete (UHPC) are combined
instead of conventional concrete and steel rebars which is widely known as “Reinforced
Concrete.” TARUHPC is intended for construction of highly durable, resilient, and sustainableinfrastructure. TiABs offer higher strength, superior fatigue performance, high strength-toweight ratio, lighter weight, lower modulus of elasticity, reduce rebar congestion, smaller
inelastic residual deformation, and excellent corrosion resistance compared to traditional
reinforcing bars. At the same time, UHPC offers higher compressive strength, superiormechanical properties, and excellent durability compared to conventional or high-strength
concrete. UHPC utilizes some of the byproducts such as fly ash and silica fume as supplementary
cementitious materials to reduce the portion of hydraulic cement in the mix. This translate in
reducing the carbon dioxide (CO2) emissions of concrete while providing excellent durability
and strength. In this research, experimental work is carried out to investigate various mechanicalproperties of TARUHPC, including bond strength of TiABs in UHPC through a series of pullout tests.
By using these test results, and previous pull-out tests, an explainable data-driven model,
based on ensemble learning, is proposed for the bond strength prediction of TiABs in UHPC.
The proposed model showed great accuracy and provided important insights for planning future
tests on TARUHPC elements. The results also showed that state-of-the-art techniques such asmachine learning in combination with TARUHPC can help to better design highly durable
infrastructure.
1. Introduction
Composite advanced novel materials have been continuously introduced in civil and criticalinfrastructures lately that would improve structural performance and durability. Some of thesemajor advanced materials that has huge potential to benefit civil and critical infrastructure
include Ultra-High Performance Concrete (UHPC) and Titanium Alloy Bars (TiABs). UHPC is
one of the advanced cementitious composite materials with exceptionally high strength and high196

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durability compared to the conventional normal concrete. One of the unique features of UHPC isthat it utilizes recycled products such as fly ash and silica fume as supplementary cementitious
materials to reduce the portion of hydraulic cement (up to 70%) in the mix. Also, the addition of
the steel fibers in the UHPC concrete can result in the added ductility in structures. UHPC offers
higher compressive strength (9 times higher), superior mechanical properties, and excellent
durability compared to conventional or high-strength concrete (Khadka et al. 2022). Similarly,
the permeability for the UHPC concrete is about 100 times less than that of normal concrete,
which is an important measure looking at the durability of the concrete.
Similarly, Titanium Alloy Bars (TiABs) is an advanced material that has been gaining
popularity for applications in concrete structures. TiABs (Ti6Al4V) offer higher strength (2.5times higher), superior fatigue performance, high strength-to-weight ratio, lighter weight (60%less), lower modulus of elasticity, reduce rebar congestion (50% less), smaller inelastic residual
deformation (50% less), and excellent corrosion resistance compared to conventional rebars(Acharya et al. 2022 and Thapa et al. 2022). It has great potential mostly in construction andretrofitting of bridges and beams. Researchers at Idaho State University (ISU) are among the
one’s to carry out several experimental and analytical research to determine the mechanicalproperties of TiABs (Acharya et al. 2021 and Khadka et al. 2020). Research was carried out toidentify mechanical properties for TiABs and compare them with a high strength alloy (e.g.,
grade 150 ksi). The grade of titanium used in the research is Grade-5 (Ti6Al4V) which is alloy ofTitanium, Aluminum, and Vanadium. A series of tests such as tension test, hardness test, Charpy
V-notch, galling tests and bond tests were carried out to compare TiABs against the 150 ksi highstrength steel alloy and clearly, TiABs performed better.
2. Pullout-Test of TiABs in Concrete
A full-scale experimental investigation was carried out to evaluate the bond-slip relationshipbetween rebar (TiABs) and concrete (UHPC and/or Normal Concrete). The size of the pullout
specimen was 22 in x 10 in x 4 in. Embedment length of 5 in, and 9 in were used for the
specimens. The specimens were tested in tension using the displacement-controlled servohydraulic actuator. The test-setup prepared for the specimen is presented in Fig. 1.

Fig. 1. Test setup: (a) Front view; (b) Side view

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Testing was carried out on a total of four specimens (See Table 1). Specimen ID denotes
the type of concrete, size of bar and the embedment length. E.g., U#6-5 denotes the specimen as
UHPC beam embedded with #6 TiABs (ø 0.75 inch) and has an embedment length of 5 inch.
Results showed better performance of the UHPC beam compared to the Normal Concrete beam
as shown in Table 1. For the test-setup provided, the Normal Concrete beams had a premature
failure with an average ultimate force of about 5.2 kips and a resulting shear stress of 0.35 ksi.
However, UHPC beams performed exceedingly better with an average ultimate force of about 32
kips and a resulting shear stress of 2 ksi.
Table 1. Pullout test results

Specimen ID
U#6-5
N#6-5
U#6-9
N#6-9
27.9925.642
35.9234.757
Ultimate Force (kip)0.025100.00266
0.05025
0.00333
Slip at Ultimate Force(inch)
55
99
Embedment Length (inch)
2.37600.47891.69400.2243*Shear Stress (ksi)*Shear Stress (τ = P/πd ); P = Ultimate force, d = Diameter of bar, = Embedment length

3. Bond Strength Prediction of TiABs on UHPCThe behavior of reinforcement, i.e., Titanium Alloy Bars (TiABs) in UHPC is generally
determined by conducting pull-out tests. However, these tests need amounts of time and cost. So,
the paper presents the application of artificial intelligence technique, i.e., machine learning, inpredicting ultimate bond strength between UHPC and TiABs in-order-to identify appropriatetechnique that can save significant amount of time and cost of experimental testing. A total of
334 experimental data from the available literature are collected. The dataset description alongwith statistical feature is presented in Table 2.
Table 2. Statistical features of the dataset
Description
Compressive Strength
Yield Strength
Tensile Strength
Bar Diameter
Embedment Length
Concrete Cover
Output Bond Strength
InputsSymbolsf’c (psi)fy (psi)fu (psi)d (in)(in)c (in)τ (psi)

MinimumMaximum
3481-29733
53896-140000
84557-150000
0.31-0.98
0.55-11.93
0.63-7.64
259-12022

Mean
19190
81033
103220
0.62
2.96
3.16
4683

Standard
Deviation
8150
18863
17074
0.15
2.26
1.64
3067

Various ML models were identified that could be built to train the pullout testing datasets.
The major models that were considered for use in bond strength prediction were: a) Random
Forest (RF), b) XGBoost (XGB), and c) AdaBoostR2 (ADB). Random Forest (Ho 1995) is thefunction that uses decision trees for regression analysis where each tree is built from a random
subset of the training set. Similarly, XGBoost (Chen et al. 2016) is an implementation of gradientboosted decision trees designed for speed, performance, and an iterative decision tree algorithmwith multiple decision trees. An AdaBoost regressor (Schapire 2013) is a meta-estimator thatbegins by fitting a regressor on the original dataset and then fits additional copies of the same198

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dataset but where the weights of instances are adjusted according to the error of the current
prediction. As such, subsequent regressors focus more on difficult cases. This class implements
the algorithm known as AdaBoostR2 which is a supervised instances-based domain adaptation
method suited for regression tasks.
The dataset is split into 80% of training and 20% of testing data. The results from the MLmodels showed that RF has the testing accuracy, training accuracy, and R2 score of 22.3%, 67%
and 0.89 respectively. Similarly, XGB was successful in obtaining testing accuracy, training
accuracy and R2 score of 92.2%, 98.2% and 0.92 respectively. AdaBoost outperformed othermodels with testing accuracy (92.8%), training accuracy (98.2%) and R2 score (0.93). This isalso verified with the Taylor diagram (Fig. 2). The results obtained with AdaBoost showed toperform better among the three models that were considered. From the result, it is evident thattools such as machine learning can be used to predict the bond strength of TiABs on UHPC. The
testing accuracy is relatively lower compared to the training accuracy on all the models; but it isclear that these tools have potential to be used for prediction.

Fig. 2. Taylor diagram
4. Conclusions
Titanium Alloy Reinforced Ultra-High Performance Concrete (TARUHPC) has a huge potential
for construction of highly durable (100+ years of service life), resilient, and sustainableinfrastructure. Research was carried out to investigate the bond-slip relationship between TiABs
reinforcing and UHPC. The results are compared against the Normal Concrete beam reinforced
with TiABs. Results showed that TARUHPC beams performed exceedingly better with anaverage ultimate force of about 32 kips and a resulting shear stress of 2 ksi. Since, TARUHPC isan expensive material to carry out significant amount of tests; an explainable data-driven model,
based on ensemble learning, is proposed for the bond strength prediction of TiABs in UHPC.
Several machine learning models were considered and AdaBoost showed the best performance(R2 = 0.93) for predicting the bond strength of TiABs on UHPC. Utilizing transfer learning,
performance comparison of different transfer learning techniques, optimization, comparison with
empirical models, and interpretability of the model is still on-going research to accurately define
the bond-slip relationship between TiABs and UHPC.

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5. References
Acharya, M. and Mashal, M. 2022. Titanium Alloy Bars for Construction of Resilient andDurable Concrete Structures. 12th National Conference on Earthquake Engineering, Salt Lake
City, UT, United States.
Acharya, M., Khadka, R. and Mashal, M. 2021. Preliminary Bond Testing and Splicing ofTitanium Alloy Bars. Transportation Research Record, 2676(5), 410-427.
https://doi.org/10.1177/03611981211067785.
Chen, T. & Guestrin, C., 2016. XGBoost: A Scalable Tree Boosting System. In Proceedings ofthe 22nd ACM SIGKDD International Conference on Knowledge Discovery and Data Mining.
New York, NY, USA: ACM, pp. 785–794. Available at:
http://doi.acm.org/10.1145/2939672.2939785.
Ho, T.K., 1995. Random decision forests. In Proceedings of 3rd international conference on
document analysis and recognition. pp. 278–282.
Khadka, R., Acharya, M., Mashal, M. and LaBrier, D. 2022.Visualization of Macroscopic
Structure of Concrete Based on X-ray Computed Tomography using Immersive Environments.
HCII 2022, Springer, Cham. https://doi.org/10.1007/978-3-031-05939-1_2.
Khadka, R., Mashal, M., and Cantrell, J. 2020. Experimental Investigation on Mechanical
Properties of Titanium Alloy Bars: Comparison with High-Strength Steel. Special Publication ofAmerican Concrete Institute. ACI Symposium Publication Vol 341 (8). 160-187.
Schapire, R.E., 2013. Explaining adaboost. In Empirical inference. Springer, pp. 37–52.
Thapa, A., Mashal, M. and Acharya, M. 2022. Large-Scale Flexural Testing of Concrete Beams
Reinforced with Conventional Steel and Titanium Alloy Bars. International Association for
Bridges and Structural Engineering (IABSE) Symposium, Prague, Czech Republic.

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Design of Ultra-High Performance Concrete Bridges
Benjamin Graybeal1* and Rafic Helou2
1
2: U.S. Federal Highway Administration, Washington, DC, USA; benjamin.graybeal@dot.gov: U.S. Federal Highway Administration, Washington, DC, USA; rafic.helou@dot.gov*: corresponding author

Keywords: ultra-high performance concrete; UHPC; codes and specifications; structural design;
engineering mechanics
1. Introduction
Advancements in concrete technology have led to a new structural material becoming available
for use in bridge design, construction, and rehabilitation — ultra-high performance concrete(UHPC). As with any novel structural material, the new standards, specifications, and guidancecovering the use of the material need to be developed so that engineering practice can safely andappropriately engage whatever advantages are associated with the new material. In the bridge
sector in the United States, there is a groundswell of interest in potential structural applicationsof UHPC, particularly those which engage the enhanced mechanical and durability properties of
UHPC to create longer span, lighter weight, resilient, and exceptionally durable structures. The
U.S. Federal Highway Administration’s UHPC research and development program has been at
the forefront of developing needed structural design guidance and engineering property test
methods, leading to a series of notable advancements that will facilitate the use of UHPC in the
U.S. bridge sector. Central to this effort is the publication of report titled Structural Design with
Ultra-High Performance Concrete that includes draft structural design guidance, draft material
conformance guidance, and a pair of structural design and analysis examples (Graybeal andHelou 2023).
2. Material Performance and Conformance
The distinct material properties of UHPC both facilitate novel applications and necessitate theengagement of new test methods and conformance requirements. UHPC-class materials provide
mechanical resistance that surpasses that expected from conventional and high performance
concretes. The tensile mechanical behavior of UHPC is a defining characteristic of the material,with UHPCs exhibiting a cracking strength greater than 5 MPa and a sustained or increasing
post-cracking strength through a localization strain of at least 0.0025. The localization strain is
the elongation at which the material ceases to display a response with multiple, tightly-spaced
cracks and begins to exhibit excessive widening of a single, dominant crack. Reliance on thetensile response of UHPC within the design of structural components necessitates the reliance on
test methods that assess and verify the assumed behaviors. To this end, a test method akin to thatused for structural metals has been developed, refined, and standardized in the United States(Graybeal and Baby 2013, AASHTO 2022). The AASHTO T 397 Standard Method of Test for
Uniaxial Tensile Response of Ultra-High Performance Concrete provides the testing and201

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analysis methodologies required for the direct determination of key engineering properties
necessary for structural design.
As with any structural material being used in an engineered system, the engineering properties
that are relied upon for system performance must be verified as being available. This verificationis commonly completed through material conformance specifications which delineate thefundamental properties that a material must exhibit, the methods for assessing the properties, andthe processes for ensuring that the delivered product is acceptable. For UHPC, key performance
metrics are the compressive mechanical response, the tensile mechanical response, and the
durability. Additional prescriptive material conformance requirements related to required
constituents, minimized contaminants, and casting and curing processes are also commonlystated. In support of AASHTO’s efforts to promulgate UHPC structural design guidance for the
US bridge sector (discussed below), the need for companion material conformance guidance isbeing addressed by the authors who have drafted a framework that is expected to be refined thenballoted by AASHTO in 2024. A working draft of the material conformance guidance can befound in Graybeal and Helou (2023).
3. Structural Design Guidance for Ultra-High Performance Concrete
Efficient and appropriate use of structural materials requires that the fundamental behaviors ofthe material are understood and engaged safely within a structural system. A first principlesbased approach, wherein engineering mechanics is heavily engaged as a predictive tool within
the design process, is a common method for ensuring that new materials can be used in disparate
applications without incurring undue risk. Although existing analysis and design methodologies
for conventional concrete can be informative, any methodologies that rely on empiricism must be
carefully considered before being engaged for a new class of concrete. To address the demandfor US-based UHPC structural design guidance, the US Federal Highway Administration is
executing a long-running program to conduct applied engineering research on UHPC structuralcomponents and to develop structural design guidance for use in the bridge sector.
During the 2018-2023 timeframe the research group completed a series of studies investigating
UHPC material behaviors (El-Helou et al. 2022, Haber et al. 2018), flexural performance (ElHelou and Graybeal 2022a), beam shear performance (El-Helou and Graybeal 2022b, 2023),
interface shear performance (Muzenski et al. 2022, 2023), and pretensioning loss behaviors(Mohebbi and Graybeal 2022, Mohebbi et al. 2022). The results of these studies, in conjunction
with research and standardized guidance developed elsewhere, were combined to create aframework for the design of UHPC structures. Through work with AASHTO’s Structural
Concrete Design Committee, a Guide Specification for Structural Design with UHPC has beendrafted and is being balloted in May 2023. Once passed and published, this document will allow
for the design of UHPC structural components within an engineering mechanics-based
framework that relies on fundamental properties of the material which have been determinedthrough direct testing methods. The FHWA-developed draft of the Guide Specification for
Structural Design with UHPC can be found in Graybeal and Helou (2023).
4. Compelling Applications
The interest in using UHPC for engineered structures is founded upon the recognition that someexisting challenges are not being appropriately addressed by existing solutions. As evidenced by202

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the perspective offered in Graybeal et al. (2020), UHPC’s novel properties can allow for thedesign and delivery of lighter, longer-spanning structures, far more durable structures, andelegant structures that facilitate combination of form and function. In the surface transportation
infrastructure sector, one such compelling application is in bridge superstructures. UHPC
elements, including pretensioned girders, post-tensioned segmental boxes, and spliced girders,
can resist greater structural demands through enhanced mechanical properties and reduced deadload, leading to longer spans and shallower structures. For example, grade separation highway
interchanges might go from including four spans to including two spans; also, highway viaducts
and approaches to long-span bridges could be constructed with longer spans and fewer piers.
Another compelling application is precast, prestressed piles. The mechanical and durability
properties of UHPC allow for the delivery and installation of piles that 1) are lighter and thuseasier to transport and handle, 2) are able to be driven more rapidly using larger equipment, and
3) are far more durable than existing steel or conventional concrete solutions. This application isparticularly relevant in coastal areas that commonly need to construct infrastructure in locations
with environmentally aggressive subsurface conditions.
Further examples include local elements within portions of structures that are subjected to harshenvironments. The use of UHPC overlays for bridge decks subjected to deicing salts is
continuing to gain adherents as the value in lengthening the lifespan of existing infrastructure isrecognized. The use of precast UHPC shells for mass concrete pours in aggressive environments
is also gaining attention as it is recognized that this solution can offer both accelerated
construction and enhanced long-term durability.
5. Concluding Remarks
As novel structural concepts centered on the advantageous mechanical and durability properties
of UHPC continue to be developed, there is a greater need for formal UHPC material
conformance and structural design guidance. The U.S. Federal Highway Administration’s
decades of work in this topic area is being leveraged by AASHTO to facilitate the drafting,balloting, and delivery of the needed guide specifications. It is anticipated that these publications
will enable owners, designers, and constructors to broadly and consistently engage UHPC within
their normal practice of delivering safe, reliable, and resilient infrastructure.
6. References
AASHTO. 2022. Standard Method of Test for Uniaxial Tensile Response of Ultra-High
Performance Concrete. T 397-22. Washington, DC: American Association of State Highway andTransportation Officials.
El-Helou, R., and B. Graybeal. 2022a. “Flexural Behavior and Design of Ultrahigh-Performance
Concrete Beams.” Journal of Structural Engineering, 148(4), doi: 10.1061/(ASCE)ST.1943541X.0003246.
El-Helou, R., and B. Graybeal. 2022b. “Shear Behavior of Ultrahigh-Performance Concrete
Pretensioned
BridgeGirders.”
Journalof
StructuralEngineering,148(4),
doi: 10.1061/(ASCE)ST.1943-541X.0003279.

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El-Helou, R., and B. Graybeal. 2023. “Shear Design of Strain-Hardening Fiber-Reinforced
Concrete Beams.” Journal of Structural Engineering, 149(2), doi: 10.1061/JSENDH.STENG11065.
El-Helou, R., Z. Haber, and B. Graybeal. 2022. “Mechanical Behavior and Design Properties of
Ultra-High Performance Concrete.” ACI Materials Journal, 119(1), 181-194,
doi: 10.14359/51734194.
Graybeal, B., and F. Baby. 2013. “Development of Direct Tension Test Method for
Ultra-High-Performance Fiber-Reinforced Concrete.” ACI Materials Journal, 110(2), 177–186.
Graybeal, B., E. Brühwiler, B.-S. Kim, F. Toutlemonde, Y. Voo, and A. Zaghi. 2020.
“International Perspective on UHPC in Bridge Engineering.” ASCE Journal of Bridge
Engineering, 25(11), doi: 10.1061/(ASCE)BE.1943-5592.0001630.
Graybeal, B., and R. Helou. 2023. Structural Design with Ultra-High Performance Concrete.
Report No. FHWA-HRT-23-077. Washington DC: Federal Highway Administration.
Haber, Z., I. De la Varga, B. Graybeal, B. Nakashoji, and R. El-Helou. 2018. Properties andBehavior of UHPC-Class Materials. Report No. FHWA-HRT-18-036. Washington, DC: Federal
Highway Administration.
Mohebbi, A., and B. Graybeal. 2022. “Prestress Loss Model for Ultra-High Performance
Concrete.” Engineering Structures, 252: 113645, doi: 10.1016/j.engstruct.2021.113645.
Mohebbi, A., B. Graybeal, and Z. Haber. 2022. “Time-Dependent Properties of
Ultrahigh-Performance Concrete: Compressive Creep and Shrinkage.” Journal of Materials in
Civil Engineering, 34(6), doi: 10.1061/(ASCE)MT.1943-5533.0004219.
Muzenski, S., Z. Haber, and B. Graybeal. 2022. “Interface Shear of Ultra-High Performance
Concrete.” ACI Structural Journal, 119(1), 267–280, doi: 10.14359/51733008.
Muzenski, S., Z. Haber, and B. Graybeal. 2023. “Monolithic and Non-Monolithic Interface Shear
Performance of Ultra-High Performance Concrete.” Engineering Structures, 281: 115667, doi:
10.1016/j.engstruct.2023.115667.

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Axial Compression Performance of UHPC Filled Square Duplex StainlessSteel Tube Short Columns
Hongyuan Tang1*, Xiaowei Hu2, Jinjun Liu3, and Duan Hong4
1: Xihua University, Chengdu, China; email: tanghyseu@163.com: Xihua University, Chengdu, China; email: 549936159@qq.com
3: Xihua University, Chengdu, China; email: 1670495023@qq.com
4: Xihua University, Chengdu, China; email: 2443515688@qq.com
2*: corresponding author

Keywords: concrete-filled steel tube; duplex stainless steel; UHPC; axial compression;
calculation model
Abstract: In order to study the axial compressive behavior, nine UHPC filled duplex stainlesssteel tube (UFSST) short columns, three concrete filled conventional carbon steel short columns,
and three hollow duplex stainless steel tube specimens are designed. The main parameters
include the thickness of duplex stainless steel tube and the strength of core concrete. The results
show that the confinement coefficient is the main factor influencing the failure mode andultimate capacity of UFSST short column. With the increase of thickness of the tube from 4 mm
to 8 mm, the failure mode of UFSST short columns changed from shear type to "waist drum"
type. And the average increase rate of ultimate bearing capacity of UFSST short columns is55.43%, the average increase rate of ductility is 17.45%. Based on the test results and confinedconcrete model of Mander, the prediction formula of ultimate bearing capacity of UFSST shortcolumn with good accuracy is established by superposition principle.
1. Introduction
Ultra-high performance concrete (UHPC) is a kind of hydraulic cementitious material with ultrahigh strength and good crack resistance under dynamic load. A lot of researches Domestic havebeen carried out on the mechanical properties of UHPC filled steel tube (UFST) columns.
However, at present, there are relatively few studies on the UHPC filled stainless steel tube
(UFSST) columns. In order to understand the mechanical properties of the UFSST with crosssection, such as the failure mode, axial load-displacement curve, and the interaction mechanism
between square stainless steel tube and UHPC, the axial compression tests of 12 UFSST withsquare cross section short columns are carried out in this paper, Taking the thickness of the outer
stainless steel tube and the strength of core concrete as the main parameters, the purpose is toprovide theoretical basis and experimental reference for the structures.
2. Experimental Program
2.1. Specimens design
The test designed square short column specimens with three thicknesses (4 mm, 6 mm and 8mm), three concrete strength grades (C40, UHPC120 and UHPC150), nominal height of 360 mm

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and a group of hollow square stainless steel tube short column specimens. The main parametersof each test specimen are shown in Table 1.
Table 1. Main parameters and test results of specimens
ttest/ mm
Btest/ mm
Ltest/ mm
Specimen number

ρs/ %

N/ kN

S-0-4

4.63

119.27360.5-

1501.6

S-0-6

6.34

119.00360.2-2742.6

S-0-8

8.32

118.04360.5-

3600.3S-40-44.42

119.28

361.5

19.52

2140.4S-40-66.28

118.57

360.2

29.37

3037.4S-40-88.53

118.04

361.0

41.40

3952.7

S-120-4

4.54

119.00

360.9

19.98

2869.8S-120-6

6.44

118.62

361.5

30.09

3429.5S-120-8

8.02

117.91

360.5

39.06

4284.6

S-150-4

4.45

119.29

360.5

19.65

2720.8

S-150-6

6.48

118.75

361.2

30.30

3423.4

S-150-8

8.29

117.71

361.5

40.20

4395.7

Note：ttest, Btest, and Ltest are the thickness, width, and height of the outer stainless steel tube, respectively.
Steel ratio ρs=As/Ac, where As and Ac are the cross-sectional area of the outer stainless steel tube and thecore concrete, respectively. N is the tested ultimate bearing capacity.

2.2. Measurement scheme and loading system
The arrangement of strain gauge is shown in Fig. 1. The data of displacement gauge, pressure
sensor and strain gauge are automatically collected by DM-YB1860 dynamic and static test andanalysis system. The loading device is shown in Fig. 2.
L/6

TT1\TT2\TT3
TL

L/3
MT1\MT2\MT3
ML

L
L/3
B/6 B/6 B/6
Displacement
meter
Strain gauge(a)The vertical view of strain gauge

BT1\BT2\BT3BL

L/6(b)Schematic diagram of strain gaugearrangement

Fig. 1. Measurement arrangement(a)loading device

Fig. 2. Loading device

3. Experiment Results and Analyses
3.1. Failure modes of specimens
The failure modes of specimens are shown in Fig. 3. The specimen is always in the collaboration
state of core concrete and stainless steel tube during the whole loading process. For the specimen
with smaller thickness, the specimen shows shear failure, as shown in Figs. 3(c) and 3(m). For

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the specimen with larger thickness, the failure of the specimen is "waist drum" type as shown in
Figs. 3(g) and 3(n).

(a)S-0-4(b)S-40-4(c)S-120-4(d)S-150-4(e)S-0-6(f)S-40-6(g)S-120-6

Crack(h)S-150-6(i)S-0-8(j)S-40-8(k)S-120-8(l)S-150-8(m)S-120-4(n)S-120-6

Fig. 3. Failure model of specimens
3.2. Axial load-displacement curve
The axial load-displacement curve of hollow stainless steel tube is shown in Fig. 4 (a). With theincrease of the thickness of the stainless steel tube, the ultimate bearing capacity is significantly
improved. Specimens S-0-6 and S-0-8 exhibit obvious strain hardening behavior, while specimen
S-0-4
has almost no strain hardening behavior.
5000

S-0-4
S-0-6
S-0-8

1000
5

10

15

20

25

∆(mm)

a
2000

c
c
S-40-4S-40-6
S-40-8

a

1000
0

b

a

0

5

10

15

20

3000

c

a

S-120-4
S-120-6
S-120-8

1000

25

0

0

5

10

15

20

a
3000 a
a

b

c

b

c

2000

S-150-4
S-150-6
S-150-8

1000

25

∆(mm)

(b)C40 specimens

c

4000

c

b

a

2000

∆(mm)

(a)Hollow steel tube specimens

b

a

b

c

N(kN)

2000

c

b

3000

b

4000

N(kN)

3000

0
0

b

4000

N(kN)

Nhollow(kN)

4000

5000

5000

5000

(c)UHPC120 specimens

0

0

5

10

15

20

25

∆(mm)

(d)UHPC150 specimens

Fig. 4. Axial load-displacement curves of specimens
Figs.4(b)~(d) are the axial load-displacement curves of specimens with ordinary strength
concrete and UHPC. The development trend of the curves can be divided into elastic stage (0-asection), elastic-plastic stage (a-b section) and plastic flow stage (b-c section).
4. Calculation of Axial Compression Bearing Capacity
4.1. Lateral effective restraint stress reduction factor
In this study, it is assumed that the effective constraint zone of the UHPC filled square duplex
stainless steel tube short columns is within (t+2At) from the outer edge of the specimen, the area
within (B-2t-2At) of the specimen edge is the weak constraint zone, B is the outer diameter of the207

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specimen, t is the wall thickness of the stainless steel tube, A is the multiple undeterminedcoefficient of the distance from the edge of the effective restraint zone to the inner wall of thestainless steel tube and the thickness of the stainless steel tube, and the boundary equation of the
weak constraint zone is assumed to be a quarter arc, see Fig. 5.
Effective
restraint area

B

𝜎𝜎𝑟𝑟

Weaklyconstrained
regionFig. 5. Core concrete effective restraint area

𝜎𝜎𝑡𝑡

𝜎𝜎𝑡𝑡

Fig. 6. Force diagram of the square stainlesssteel tube

4.2. Calculation method of axial compression bearing capacity
According to Sakino et al, under the ultimate bearing capacity state, the longitudinal stress σ l andtransverse stress σ t of steel tube is given by:

σ l = 0.89σ 0.2(1)

σ t = 0.19σ 0.2(2)

According to the stress equilibrium condition shown in Fig. 6, the constraint stress of steel pipeon core concrete under ultimate load state is as following:

σr =

2tσ t
B − 2t(3)

Based on YAN et al,the lateral effective restraint stress is determined by:(4)

σ 'r = K etσ r

According to the constrained concrete model proposed by Mander et al, the axial compressive
strength of core confined concrete is determined through Eq. (3):f cc = f ck (−1.254 + 2.254 1 + 7.94

σ r'
f ck

−2

σ r'
f ck

)

(5)

As is the cross section area of stainless steel tube, and is given by:(6)

As = 4(c − t )t208

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where, c and t are the width and thickness of the outer square stainless steel tube.
The axial compression bearing capacity of square UFSST column is calculated by thesuperposition principle, as shown in Eq. (5):(7)

N uc = f cc Ac + σ l As

5. References
Mander, J. and Priestley, M. 1988. Theoretical Stress-Strain Model for Confined Concrete,
Journal of Structural Engineering, 114(8),1804-1826.
Sakino, K., Nakahara, H., and Morino, S. 2004. Behavior of Centrally Loaded Concrete-Filled
Steel-Tube Short Columns, Journal of Structural Engineering,130(2), 180-188.
YAN , Yanxiang . , XU , Lihua . , a n d CAI , Heng. 2019 . Calculation methods of axial
bearingcapacity of short square UHPC filled high strength steel tubular columns, Journal of
Building Structures, 40(12), 128-137.

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Seismic Retrofit of Precast and Cast-in-Place Piers
Using Ultra High-Performance Concrete (UHPC)
Kathryn Hogarth1, Manish Acharya2, Arya Ebrahimpour3, and Mustafa Mashal4*
1: Idaho State University, Pocatello, Idaho, United States; email: hogakat2@isu.edu: Idaho State University, Pocatello, Idaho, United States; email: manishacharya@isu.edu
3: Idaho State University, Pocatello, Idaho, United States; email: ebraarya@isu.edu
4: Idaho State University, Pocatello, Idaho, United States; email: mashmust@isu.edu
2*: corresponding author

Keywords: Bridge Piers; Seismic Retrofit; UHPC; Concrete Jacket; Large-scale Testing
Abstract: Recently Idaho State University completed large-scale experimental and analyticalinvestigation of a new precast pier system for applications in seismic zones. The precast piersystem uses the concept for Concrete Filled Steel Tubes (CFSTs) in plastic hinge locations.
Extensive testing on precast cantilever and bent piers was conducted to investigate seismic
performance under quasi-static cyclic loads. Results were compared against benchmark cast-inplace specimens. Testing showed superior performance of the precast pier compared to cast-inplace. Based on the experimental results, the precast pier technology was implemented in an
actual bridge located in seismic zone in Idaho. For wider applications of the precast piertechnology, it is important to investigate appropriate retrofitting strategies for post-earthquake
repair. Some of the proposed methods to retrofit CFSTs have only been investigated analyticallyand have not been experimentally tested. There is a critical need for experimental results tovalidate the analytical models before any retrofitting methodologies may be implemented onsite.
In this research, concrete jacketing using Ultra High-Performance Concrete (UHPC) is proposed
as an innovative retrofitting strategy for the precast pier. UHPC offers advantages such as highstrength, excellent durability, smaller jacket thickness, accelerated construction, strong bond withhardened concrete and reinforcing bars, among other benefits. The UHPC jacket is intended tore-instate the pier’s stiffness, strength, and ductility after the earthquake. In this research, fourdamaged columns will be tested; two precast columns incorporating CFSTs in plastic hinges andtwo cast-in-place columns as benchmark specimens. The research will show if the specimens
retrofitted with UHPC jackets have comparable seismic performance to the original specimens.
1. Introduction
This project focuses on the retrofitting of concrete bridge piers. In the first phase of the project,
two half-scale bridge bents were designed and experimentally tested in the Idaho State
University (ISU) Structural Lab (SLAB) under earthquake style loading. One of the bents
modeled a typical cast-in-place (CIP) bridge connection, while the other bent modeled a pierconnection developed by ISU and Idaho Transportation Department (ITD). The design of thebridge connection incorporating CFSTs can be seen in Figure 1. Both bridge bents were tested to210

Bridge Engineering Institute Conference 2023 (BEI-2023)
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failure and the results were processed and compared to one another. After testing of the bents,
the piers have been dismantled and are to be retrofitted for testing.

Fig. 1. Pipe-Socket Connection for Column-to-Footing Connection (1’=30.4cm, and 1”=2.54cm)
2. Literature Review
A literature review of retrofitting methods for bridge piers was conducted and investigatedmethods including steel jacketing, concrete jacketing, fiber reinforced polymers (FRP), andexternal yielding elements. The most ideal retrofit method found was the use of ultra-highperformance concrete (UHPC) jacketing. Normal Strength Concrete (NSC) has lower strengthcompared to UHPC in both compressive and tensile strength. UHPC's strength spans from
124.11 to 227.5 MPa, which is significantly higher than NSC's compressive strength, which isonly between 20.7 and 41.3 MPa. Additionally, UHPC's tensile strength is higher than NSC's,with a range of 6.89-24.13 MPa compared to NSC's range of 2.75-3.45 MPa.
2.1 UHPC Jacketing Literature Review
A study by (Farzad et al., 2019) investigated the application of UHPC jacketing in bridge repair
strategies. The results indicated that this approach significantly improved the lateral strength,
deformation, energy dissipation capacity, and stiffness deterioration of the structure.
Additionally, (Joe and Moustafa, 2016) demonstrated that utilizing UHPC instead of
conventional concrete in bridge piers has several advantages. Not only does it require 75% lessmaterial, it is also better for the environment as it can lower the carbon footprint by up to 35%
when used with a compressive strength of about 172.37 MPa. Due to these benefits, the numberof UHPC bridges is growing each year. As of 2020, there were 341 in-service bridges in the
United States that use UHPC, and the trend is increasing each year (Office of Research,
Development, and Technology at the Turner-Fairbank Highway Research Center, 2021).
3. Concept for Retrofit
After literature review, the basis of retrofitting the precast pier is to utilize a UHPC jacket in theplastic hinge zone of the pier. The concept can be seen in Figure 2.

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(a)

(b)

Fig. 2. Overall Concept for Retrofit: (a) Concept for UHPC Jacket (Green); (b) Cross-section ofPrecast 1 Retrofit (1 in. = 2.54 cm)
Note that since the pipe in this specimen is fractured, the moment capacity contribution from thepipe is taken to be negligible (i.e., a conservative assumption). Also note that for this specimen
all of the cover concrete spalled, therefore the entire section, excluding the pipe, is taken to be
UHPC. The jacket is designed to sit at a 4 in. (10.16 cm) thickness out from the original column
diameter (14 in. or 35.56 cm), with an overall diameter of 22 in (55.88 cm). After the success ofPrecast 1, Precast 2 is constructed without caging, but with the jacket thickness of 4 in. (10.16cm). Following the success of Precast 1 and Precast 2 piers, CIP 1 and CIP 2 are retrofitted withsmaller jackets of 3 in. (7.62 cm) and 2 in. (5.08 cm), respectively.
4. Construction and Test Set-up
The first step for construction is to roughen the surface for the pour, as well as drill the concrete
to allow for the tension dowels to be epoxied in. The next step is to assemble the cage. (Note,
only Precast 1 has epoxied dowels and caging). Lastly, the form is assembled and the UHPC ispoured. Note that the jacket is poured while the pier is under a 30-kip (133kN) axial load tosimulate the structure of a bridge. After the pier is poured and cured, the pier is instrumented andmoved into the lab to get ready for testing. The test is accomplished by programming thehydraulic actuator to run a loading protocol similar to the bent loading protocol, which has beenscaled down for one pier. The loading protocol is a quasi-static cyclic loading.
5. Results
A summary of the testing results can be seen in Table 1. Furthermore, Figure 3 gives the
backbone curves for the piers.

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Table 1. Summary of Results
Precast 1
Precast 2
CIP 1
CIP 2
Maximum Force
30.76 kip30.65 kip
25.08 kip25.2 kip(136.83 kN)(136.34 kN)(111.56 kN)(112.1 kN)
Maximum Drift
7.38%
9.38%
9.81%
10.19%
Moment Capacity
133.3 kip-ft.
132.82 kip-ft. 108.68 kip-ft. 109.2 kip-ft.
from Top of Jacket (180.73 kN-m) (180.08 kN-m) (147.35 kN-m) (48.06 kN-m)
Initial Yield Drift
0.76%
0.95%
1.22%
1.07%
Dissipated Energy

216.5 kJ

262.7 kJ
316.08 kJ(154.1 kJ)*
(167.4 kJ)*
* corresponding energy dissipated at 7.38% drift (failure of Precast 1)

322.76 kJ(127.4 kJ)*

Fig. 3. Backbone Curves of Bridge Piers
6. Conclusion
This project covers seismic retrofit of pipe-socket connected piers. The project investigated thesuitability of UHPC jacketing methods. For Precast 1, a 4 in. (10.16 cm) UHPC jacket is usedwith light caging. Precast 2 incorporated a 4 in. (10.16 cm) UHPC jacket with no caging. Bothmethods for retrofit were successful. Precast 1 pushed the plastic hinge up the column face withno separation from the footing. Precast 2 succeeded in pushing the plastic hinge up the face ofthe column with separation from the footing and partial failure due to rocking. CIP 1
incorporated a 3 in. (7.62 cm) UHPC jacket, and CIP incorporated a 2 in. (5.08 cm) UHPC jacket.
CIP 1 and 2 were both successful in pushing the plastic hinge up the face of the column. CIP 1
had little to no gap opening, while CIP 2 had a larger gap opening, likely due to the smaller
jacket size. Large-scale testing has recently been completed at ISU, further data will be presented
and published as it is analyzed.
7. References
Farzad, M., Shafieifar, M., and Azizinamini, A. 2019. Retrofitting of bridge columns using
UHPC, Journal of Bridge Engineering, 24(12).
Joe, C.D. and Moustafa, M. 2016. Cost and ecological feasibility of using UHPC in Bridge Piers.
First International Interactive Symposium on UHPC.
Office of Research, Development, and Technology at the Turner-Fairbank Highway ResearchCenter. 2021. Deployments: Deployments of UHPC in Highway Bridge Construction.

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Railway Bridges

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Numerical Investigation of Aerodynamic Characteristics of High-Speed
Railwaytrain-Bridge System under Tornado-Like Winds
Xuhui He1 and Simin Zou2*
1: School of Civil Engineering, Central South University, Changsha, China;
National Engineering Research Center for High-Speed Railway Construction, Changsha, China
email: xuhuihe@csu.edu.cn
2: School of Civil Engineering, Central South University, Changsha, China;
National Engineering Research Center for High-Speed Railway Construction, Changsha, China;
email: simin_zou@csu.edu.cn*: corresponding author

Keywords: Tornado; High-speed railway; Trains; Bridges; Aerodynamic characteristics
Abstract: As the high-speed railway emerges in Eurasia, a comprehensive understanding of the
aerodynamic problems – particularly extreme wind events – is vital to the success of the safety,
operational efficiency, and transportation industry. In addition to very high and transient windspeed in a relatively short time, some key distinguishing attributes of the winds associated withtornadoes are the swirling winds with multiple wind directions and the probable existence of avery different wind profile. Such knowledge of the effect of extreme wind on the train and bridge
system has been hindered by a lack of available field test data. However, limited full-scale data
along the height and radial position are available to identify and reliably establish a descriptionof this wind velocity profile and field with their effect on the high-speed train. In light of limited
full-scale data to arrive at a consensus on quantifying key parameters characterizing the tornado
winds, the wind field associated with each component of tornado winds, such as the velocity
field and pressure field and propagation to load effects on the high-speed railway train-bridge
system are carried out using high-fidelity computational fluid dynamics (CFD) in this study.
1. Introduction
Compared with other transportation means, the high-speed railway has many advantages: highefficiency, energy-saving significantly and smooth operation, therefore, it has been widely
developing globally (He & Zou, 2021). Considering the high-speed train need to run in a very
coordinated, straight-line fashion sustaining locomotion for a considerable time, the bridges havebeen widely used as a vital component in the high-speed railway (HSR) network worldwide. For
example, the Japanese HSR system has an average bridge ratio of 47% and Chinese Taiwan isreported as 73%. The average bridge occupation ratio is greater than 50% and the BeijingShanghai Express railway reaches 97% in China. In addition, the frequency of severe tornado
disasters worldwide has been rising and higher and China has not been spared. According tostatistics, there has been an annual average of more than 43 tornadoes in the past 50 years,
frequently occurring in recent years with a far larger rate than previous estimates. The worst
tornado of all time occurred in 2016, an EF4 tornado affecting the city of Yancheng in Jiangsu
Province, which killed 99 people and injured more than 800 people. It should be noted that215

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almost all tornadoes occur each year in eastern China (especially in Jiangsu and Guangdong), but
the location of a single tornado does not significantly affect laws. However, there is a high riskof railway exposure to tornadoes amid global warming.
Numerical simulations have a critical role in studying tornadoes' near-surface wind fields.
Numerical simulation is more efficient and economical to analyze the three-dimensional
velocities and pressure fields due to the strong turbulence motion near the surface and the centerof the vortex. Especially numerical simulations can generate complete and detailed data in space
and time, which allow for calculations of time-averaged wind speeds, short-term gusts, flow
fields, and variations in space and time. Such information is particularly vital to the civilengineering community. With the rapid development in recent years, fast convergence rate, low
computational complexity, and excellent numerical stability of the numerical algorithm are thegoals we continuously strive to pursue.
In this study, the tornado-induced effect on a high-speed train running on a bridge wascalculated numerically by IDDES simulations in a three-dimensional model. In Section 2, thenumerical algorithms used in this study were introduced. The details of the tornado simulationand calculation accuracy were discussed in Section 3. The tornado-induced aerodynamic forces
acting on a high-speed train were investigated in Section 4. In order to decrease the accident rateof the train and evaluate the aerodynamic performance, the current common height of the train in
operation is created and analyzed in this article.
2. Numerical Algorithms
All the simulations presented were completed in the commercial software Ansys Fluent
19.2. The time-dependent IDDES (based on the SST k-w model) used in this paper is a hybrid
RANS-LES model. The difficulties associated with using the standard LES models, particularly
in near-wall regions, have led to the development of hybrid models that combine the best aspectsof RANS and LES methodologies in a single solution strategy (Spalart et al.,1997).
In the DES approach, the unsteady RANS models are employed in the boundary layer,
while the LES treatment is applied to the separated regions. The LES region is normallyassociated with the turbulent core region, where large unsteady turbulence scales dominate. In
this region, the DES models recover LES-like subgrid models. In the near-wall region, therespective RANS models are recovered.
The Improved Delayed Detached Eddy Simulation (IDDES) model is a hybrid RANS-LES
model (consisting of a combination of various new and existing techniques) that provides a more
flexible and convenient scale-resolving simulation (SRS) model for high Reynolds number flows.
3. Tornado Generation
The regression equations of tangential velocity and radial velocity along the height are expressed
as follows:𝑧𝑧

𝑉𝑉𝑡𝑡 = 20.61(20)0.1774(4)

𝑉𝑉𝑟𝑟 = �(5)𝑧𝑧– 31.14(20)0.169 , 𝑧𝑧 < 20𝑚𝑚𝑧𝑧

45.14(20)0.1826 – 76.48, 𝑧𝑧 ≥ 20𝑚𝑚

where z is the height above the ground, these wind velocity profiles are used as the input at thevelocity-inlet within the inlet height range.
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Fig. 1 shows the mesh system of the numerical tornado simulation. In order to accurately capture
the flow fields of tornado-like vortices and quantitatively investigate the wind loading onthe train
in the central part of the convergent zone and the vicinity near the ground, a fine mesh isconsidered.
b）

(a)

(b)

1.0

1.0

0.8

0.8

0.6

0.6

V0

V0

Fig.1 Computational domain and grid layout

Rankine model

0.4

Rankine model

0.4

Rankine modified model x=0.4

Rankine modified model x=
Rankine modified model x=0.5

Rankine modified model x=0.5
Rankine modified model x=0.6

0.2

Rankine modified model x=0.6

0.2

Spencer z=0.1R

Spencer z=0.52R
Mulhall z=0.52R

Mulhall z=0.1R

0.0

Simulation

0

1

2

r/R

3

0.0

4

Simulation

0

1

2

r/R

3

4

Fig.2 Comparison of the tangential velocity distribution with those of the tornado and model at
different heights (Z=0.1R & Z=0.52R)
Fig.2 presents the normalized tangential velocity profiles to compare with those of the tornado
events and the analytical model. The distance from the vortex center, r, is normalized by the coreradius rm . The velocity of tornado events from full-scale tornados at Spencer, South Dakota, andMulhall, Oklahoma, recorded by Haan et al.(2008), are plotted with the present
result in the figure. For the Rankine model, the overall maximum tangential velocity VR of eachvolume and the corresponding radius rm were used to calculate the tangential velocity in thesemodels. Overall, the simulation result better agrees with the radar data and analytical model.

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Also, decay factors of 0.4-0.6 result in a better match with the simulation, consistent withprevious studies findings.
4. Tornado Effects on a Running Train on the Bridge
6

CFX
CFY
CFZ

30

Core

15
train leaving
the tornado

train entering
the tornado

0

−15

−30

Aerodynamic force coefficents

Aerodynamic force coefficents

45

−45
0

20

40

60

80

100

120

140

Time (s)

CFX
CFY
CFZ

4

2

Core
train leaving
the tornado

train entering
the tornado

0

−2

−4

−6

0

5

10

15

20

25

Time (s)Fig. 3. Time history of aerodynamic forces coefficients
The model train and bridge were made based on a 1:1 scale of the CRH train with three cars and asimply-supported box girder. The bridge is selected to locate about 4.5 m above the ground andalong the whole diameter of the tornado. In order to get the historical time forces on the train, thetrain was tested in a stationary tornado with a train operation speed of 50 km/h and 300km/h. At
each location, the time-averaged aerodynamic forces were obtained by simulating the 400s, andthe first 10s data were removed to eliminate the transient results. Because of the almost zero
tangential velocity at the center of a tornado, the resulting aerodynamic force distribution here
does not vary substantially. The absolute values of aerodynamic force coefficients go for a
maximum at the core region due to the effect of the tornado’s maximum tangential velocity,
indicating that the tornado flow has a devastating effect. The large value of CFz at the r/R=1 andbeyond should be due to the large vertical wind velocity near its core boundary. It means aground effect exists when the train and bridge are near the surface and buoyancy on the bridge
falls off with the distance from the tornado center. It should be noted that one of the main
damage from tornadoes is caused by buoyancy.
5. Conclusion
Improved delayed detached eddy simulation (IDDES) for tornado ﬂow around a train wasapplied and the results show Aerodynamic forces coefficient of sections varies with distancefrom the tornado center, the absolute values reach a maximum at r/R=1. Meanwhile, enormous
buoyancy is generated here due to a ground effect exists.This was the ﬁrst step toward a realistic
simulation that reproduces the effect of a tornado on a running train on the bridge. It should be
noted that the train at the different positions in the tornado is still being concerned, more details
on the effects of the tornado will be explored further.
6. References
Haan, F.L., Sarkar, P.P., Gallus, W.A. Design, construction and performance of a large tornado
simulator for wind engineering applications. Eng.Struct. 2008, 30(4),1146-1159.
He X.H, Zou S.M. Advances in wind tunnel experimental investigations of train–bridge systems.
Tunnelling and Underground Space Technology, 2021,104157.
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Effects of Splitter Plates on the VIV Performances of Twin Separated
Parallel Decks for a Rail-Cum-Road Bridge
Lulu Liu1, Xuhui He2*, Yunfeng Zou3, and Zhen Wang4
1: Central South University, Changsha, China; email: liululu2020@csu.edu.cn: Central South University, Changsha, China; email: xuhuihe@csu.edu.cn
3: Central South University, Changsha, China; email: yunfengzou@csu.edu.cn
4: Central South University, Changsha, China; email: wangzhen@csu.edu.cn*: corresponding author
2

Keywords: Splitter plates; Vortex-induced vibration (VIV) ; Wind tunnel test; The rail-cum-road
bridges; Computational Fluid Dynamics (CFD)
Abstract: Installing the splitter plates at decks is among the passive aerodynamic solutions foreliminating vortex-induced vibration (VIV). However, the influence of splitter plates on the VIV
performances are more complicated due to aerodynamic interference between highway andrailway decks. To study the effects of splitter plates, wind tunnel experiments were conducted formeasuring VIV performance under two opposite flow directions, while the surrounding flow
field of static twin decks with and without splitter plates is numerically simulated byComputational Fluid Dynamics (CFD). A wide range of width (L) varying from 0 to 80 mm fortwin asymmetrical parallel models are selected to investigate the effect of splitter plates on VIV
performance The results showed that the incoming flow direction affects the VIV response. The
highway deck has poor vertical and torsional VIV performance, and the VIV region andamplitude are different under differ locations. While the railway deck only has vertical VIV
when located upstream. In addition, the splitter plates can impede the process of vortex
generation, shedding and impinging at the gap between the highway and railway deck, thus
effectively suppressing the VIV of twin decks.
1. Introduction
The Long span twin parallel bridges are a unique type of rail-cum-road bridges that may greatly
increase cross-straits' capacity and enhance transportation. However, wind-induced response
problems such as static wind stability and interactive vortex-induced vibration (VIV) of bridges
caused by complex inter-gap aerodynamic disturbance effects (Seo et al., 2013). The thin-walled
splitter plate is one of the effective measures to control the VIV among passive controls (Akilli etal., 2005). There is little relevant research about the influence of the thin-walled splitter plate on
the VIV performance, and the research on splitter plate is always for a single bridge but an
integrated evaluation of the overall performance of twin parallel decks. Is this method of splitter
plates on VIV-suppression applicable, or are the consequences of the VIV-suppression effect andaerodynamic characteristics on the adjacent bridge more complex? All these issues need to beconsidered in the study. Therefore, the effect of splitter plate on wind-resistance performance oftwin separated parallel decks requires further detailed investigation.

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2 Experimental Study of Vortex-Induced Vibration
2.1 Experimental setup
Fig. 1 graphically depicts the span deck cross section. As shown in Figs. 2 and 3, using an
independently constructed twin decks model elastic suspension device, the VIV measurement
test was carried out at the wind tunnel. Four laser displacement transducers with 500 Hz
sampling rates and 30 s sampling times were used to measure the vibration response (see Fig. 2).
Vertical and torsional vibration could occur freely in the highway and railway model. The
dynamic features of the highway and railway bridges are shown in Table 1. Seven different
widths of splitter plates were made from ABS plates and fixed to each side of the twin decks inturn in order to study the VIV-suppress effect of the splitter plates. Wind tunnel experiments
were conducted to test the VIV amplitude of the decks under various wind directions. The exact
test conditions are shown in Table 2.

Fig. 1. Cross section at mid-span position (unit: m)

Splitter plate

Fig. 2. System for section model testing support
Table 1. Dynamic parameters
Parameter
Geometric scale λs
Mass m (kg/m)
Mass moment of inertia J(kg∙m2/m)
Vertical frequency fv (Hz)Torsional frequency ft (Hz)Fig. 3. VIV tests of wind tunnel

Railway decks
bridge
model
1:50
49600
16.28

Highway decks
bridge
model
1:50
52500
21

2870000

0.452

2350000

0.376

0.463
1.031

5.371
11.963

0.348
0.856

4.028
9.930

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Table 2. Test conditions
Wind
Cases
directions
Case1
Figure 5 (a)Wind angle α3°Damping ratio
ζv/ζt
0.34%/0.32%

Width of splitter plates (scalemodel) L (mm)0, 20, 30, 40, 50, 60, 70, 80

2.2 VIV Performance with Various Splitter Plates
As shown in Fig. 4(a), when the wind blows from the UH to the DR, the vertical VIV
displacement of the UH deck gradually decreases with the width of the splitter plate. When the
splitter plate width is L = 50 mm, one of the VIV regions even disappears directly, the vertical
VIV region becomes only one, and the maximum VIV amplitude is 82 mm. When the width
increases to L = 80 mm, the vertical VIV of the UH deck does not occur at all. As shown in Fig.
4(b), for the torsional VIV, the VIV response gradually becomes weaker as the splitter plate's
width increases. When the width is L = 20 mm, the torsional VIV region is changed from three toone, and the maximum torsional amplitude is reduced by 20%. As the width of the splitter plates
keeping increasing, the amplitude of the torsional VIV continues to decrease, and the speed
range continues to decrease. When the width is L = 50 mm, the torsional VIV for UH deck
completely disappears. And if the width of the splitter plate continues to increase, there is also no
torsional VIV. It demonstrates that the splitter plate can successfully reduce the torsional VIV.

(a) The vertical displacement(b) The torsional displacement

Fig. 4. VIV of the twin decks (Wind blows from UH to DR)
As shown in Fig. 5(a), when the wind blows from the UR to the DH, the vertical VIV
displacements and VIV region of the UR and the DH deck also gradually decrease with the
splitter plate's width. When the width is L = 20 mm, the DH deck's maximum vertical VIV
displacement is decreased to 82 mm, which is reduced by 73%, which also meets therequirements of the specification. When the width increases to L = 50 mm, the VIV of the UR
even disappears directly. The DH deck's vertical VIV amplitude is sensitive to the splitter plate's
width. But for the DH decks, As the width of the splitter plate keeping increasing, although theamplitude of the vertical VIV continues to decrease, the amplitude is obviously smaller, but thevertical VIV does not completely disappear until the width is L = 80 mm. As shown in Fig. 5(b),for the torsional VIV, with the increase of the splitter plate's width, there is no new torsional VIV
phenomenon in the DR deck.

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(a) The vertical displacement(b) The torsional displacement

Fig. 5. VIV of the twin decks (Wind blows from UR to DH)
To sum up, the splitter plates on each sides of the twin decks affects the vertical andtorsional VIV. It is more appropriate to comprehensively select the width for L = 50 mm.
3. Numerical Simulation of Flow Field around Static Decks
2D models of static highway and railway decks was chosen for CFD simulation to examine theinfluence of splitter plates on the flow visualization. On the basic of the wind tunnel tests
parameters, the computational domain settings in the simulations made are shown in Fig. 14.
taking the L = 50 cm splitter plates as an example, the numerical simulation results are comparedwith and without the splitter plates on twin decks to analyze the effect of the splitter plates.

Fig. 6. Layout of computation domain discretization and mesh scheme
As shown in Fig. 7(a), distinct vortices periodically shed from the leeward side of the upperand lower surfaces of the UR deck and then impinged on the windward side of the DH deck. The
vortex shed from the UR deck' upper surface flows directly over the upper surface of the DH
deck, while the vortex shed from the UR deck' lower surface impinged the winding nozzle of the
DH deck. And when the splitter plates were installed on each sides of the twin deck, the splitter
plates make the flow field separate in advance and affects its wake shedding. Besides, as shownin Fig. 7(b), the flow around the railway and highway decks becomes smooth and no periodic
vortices are generated. In other words, once the splitter plates are installed, the highway andrailway bridges tends to be streamlined under the airflow.

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Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023(a) Without splitter plates (b) With splitter plates
Fig. 7. Vortex contour around twin girders
4. Conclusion
The splitter plates can impede the process of vortex generation, shedding and impinging at the
gap between the highway and railway decks, thus effectively suppressing the VIV. So that theincoming flow can flow steadily through the railway deck upstream and the highway deck
downstream.
5. References
Akilli, H., Sahin, B., Filiz Tumen, N., 2005. Suppression of vortex shedding of circular cylinder
in shallow water by a splitter plate. Flow Measurement and Instrumentation 16, 211-219.
Seo, J.-W., Kim, H.-K., Park, J., Kim, K.-T., Kim, G.-N., 2013. Interference effect on vortexinduced vibration in a parallel twin cable-stayed bridge. Journal of Wind Engineering andIndustrial Aerodynamics 116, 7-20

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Aerodynamics of a High-Speed Train Crossing the Wake of a Bridge Tower in
Different Yaw Angles
Zuyu Xie 1,2, Huan Li 1,2*, Xuhui He 1,2 , and Zhirui Huang 1,2
1: School of Civil Engineering, Central South University, Changsha, China; LI Huanlihuan2016@csu.edu.cn
2: National Engineering Research Center of High-speed Railway Construction Technology,
Changsha, China
Abstract: Bridge tower shows different effects obviously in different yaw angles for its
windward area varies, which may result in complex wake and cause transportation problems (Wuet al. 2022). This study concentrates on the aerodynamic characteristics of a CRH-3 high-speed
train passing through the bridge tower at the yaw angle of 15° and 90° . Shielding Effect of thetower bridge between the yaw angle of 0° ~ 90° at the location of the peak value of CFi. is alsostudied The scaled ratio of the train, bridge and bridge tower are 1:25. In this test, the train isfixed on the bridge and the bridge tower is placed on guideways. Aerodynamic force of the train
is tested with the change of location of the bridge tower. The experiment is tested in a close-loop
low-speed wind tunnel. The test section is 12m in width, 3.5m in height, and 18m in length. The
wind speed in the low-speed section can vary from 0 to 25 m/s, with the turbulence intensity
lower than 0.5%.The test model is shown in Fig.1.

Fig. 1. Layout of test
For the abstract, the results of leeward condition are mainly described.

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Fig. 2. Aerodynamic force of CRH-3 trains (β = 15°)

Fig. 3. Aerodynamic force of CRH-3 trains in different yaw angles (β = 0 ~ 90°)As the Fig.2 shows, in the leeward condition, the aerodynamic forces on the head train, middle
train, and trailing train were significantly affected by the wind-shielding effect of the bridge
towers, with significant changes observed in the CFx and CFm coefficients. When each of thethree train was blocked by the bridge towers, there was a sudden increase in aerodynamic forces,
with a noticeable sinusoidal variation pattern. In Fig.3, as the yaw angle changed, the bridge
towers significantly affected the vertical aerodynamic force coefficient, CFz, of the leading car.
Within a yaw angle range of 0° to 40°, the variation range of CFz for the leading car varies
slightly, which indicates the head train is affected by the bridge tower, which is consistent withthe actual situation.
1. References
Wu, J. F., et al. "Aerodynamic characteristics of a high-speed train crossing the wake of abridge tower from moving model experiments." Railway Engineering Science 30(2): 221-241,(2022).

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Concrete Structure Damage Analysis of the Anchoring End of Wind Barriersof High-Speed Railway Bridge under Wind Load of Train
Haobo Liang1, Yunfeng Zou2*, Xuhui He3, and Dianyi Guo4
1

: School of Civil Engineering,haoboliang0816@163.com
2: School of Civil Engineering,
yunfengzou@csu.edu.cn
3: School of Civil Engineering,
xuhuihe@csu.edu.cn
4: School of Civil Engineering,
guodianyi@csu.edu.cn

Central South University, Changsha, China; email:Central South University, Changsha, China; email:Central South University, Changsha, China; email:Central South University, Changsha, China; email:*: corresponding author

Keywords: concrete damage; wind barriers; high-speed railway bridges; train wind load; concrete
damaged plasticity model
Abstract: The installation of wind barriers can effectively reduce the impact of cross wind on the
train and improve the safety and comfort of the train running on the bridge. However, a largetransient aerodynamic pressure will be generated when the train passes through wind barriers at
high speed. Repeated impact of such pressure wave on the wind barrier for a long time may cause
damage and cracking of the concrete at the anchor end of the wind barrier, which will seriously
threaten the operation safety of the high-speed train. In this study, based on computational fluid
dynamics (CFD) analysis, the time history curve of train wind load acting on the wind barriercolumn is obtained. Then, ANASY is used to calculate the displacement time-history curve of thewind barrier column under the train wind load, and determine the most unfavorable position of theload on the wind barrier column. Finally, the finite element model of the wind barrier columnconcrete pedestal was established by ABAQUS to explore the stress and damage distribution of
concrete structure at the anchoring end of wind barrier under train wind load. The research results
serve as a reference for the safety assessment of damage and its maintenance and reinforcement of
concrete structures at the anchoring end of wind barriers.
1. Introduction
Wind barriers can effectively improve the safety of train running in windy environment, but its
structure needs to withstand the pulsating wind pressure generated by the passage of high-speed
trains, which will threaten the structural safety of the wind barrier. The expansion and compression
effects of the train pulsating wind pressure have caused vibration, fatigue and loosening of bolts
at the connection site of the wind barrier components (Xiong et al., 2020). As the speed of the train
increases from 300km/h to 400km/h, the pulsating pressure caused by the train will increase
rapidly with the square of the train speed (Tian et al., 2007), and the corresponding safety problemof wind barrier structure will become more prominent. Moreover, the concrete fatigue effect at theconnection between the wind barrier column and the bridge structure is small. The concrete at theconnection of the column foundation is prone to crack under the action of fatigue working load.

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At the same time, according to the test results of scholar (SOPER D et al., 2019): the maximum
lateral force under the pulsating wind pressure load of the train (running speed of about 228km/h)
can reach 2166N, the load cycle of a single running direction can reach 234 times per day, and thenumber of load cycles in the 20-year maintenance-free period exceeds 1.5 million. With thecontinuous action of wind load on the wind barrier, the concrete cracks will increase and deepen,
and the consequent cumulative damage, fatigue effects and changes in the mode of stress, etc., if
not dealt with in time, it is highly likely lead to serious wind barrier overturning. Therefore, it isnecessary to study the effect of the pulsating pressure generated when the train passes through on
the damage of the concrete structure at the anchored end of the wind barrier.
2. Calculation Method
2.1. CFD numerical simulation
ANSYS Fluent was used to establish a numerical simulation model of wind barrier-train-bridge,
which used a typical high-speed railway 32m double-line simply supported girder bridge andCRH2 Harmony high-speed train as the research objects, and used a wind barrier with a height of
3m and wind permeability of 0%. The high-speed train adopts a three-car formation model (front,
middle and tail) with a total length of 76.4m (25.7+25+25.7m). The total length of the bridge modelis 352m, and the wind barrier is set in the middle of the entire length of the bridge. The threedimensional compressible SST k-ω turbulence model was selected for the calculation model, andthe 3-marshalling CRH2 bullet train was simulated at the speed of 400km/h through a 352m longbridge by sliding grid method, and finally the wind load time of the train was obtained, as shownin Fig. 1.

Fig. 1. Comparison between measured and simulated aerodynamic impulse pressure
2.2. Structural dynamic response analysis
Based on the actual material parameters of the wind barrier column, the design parameters were
defined and the ANSYS finite element model was established. The train wind load obtained by CFD
numerical simulation is applied to the wind barrier column. Meanwhile, considering the travelingwave effect of the train wind, the dynamic response of the wind barrier column under the train
pulsating wind load is calculated by transient analysis method, and the maximum displacement
time history curve of the wind barrier column is calculated, as shown in Fig. 2.

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2.0

Displacement/mm

1.5
1.0
0.5
0.0
-0.5
-1.0

H-shaped column top node
(Node-67533)

-1.5
-2.0

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

Time/sFig. 2. Maximum displacement response time history of wind barrier column
2.3. Structural nonlinear finite element damage analysis
A finite element model consisting of wind barrier column, concrete pedestal, and U-shaped anchor
bolts (M27) was established using ABAQUS, as shown in Fig. 3. Based on the code provisions,apply sufficient pretightening force (calculation formula of bolt pretightening force: FT=T/K*D,
where T is the torsional torque; K is the dimensionless coefficient of torsional torque, 0.22; D isthe nominal diameter of the bolt, 27mm.) to the U-bolt to compress the wind barrier column andconcrete pedestal, where the interaction between the wind barrier column and concrete is set as
surface-to-surface contact. The analysis step is set up in two steps, the first step considers only thebolt pretightening force, and the second step adds the train wind load. The nonlinear constitutive
relationship of concrete is defined by the plastic damage model (CDP) derived based on thespecification. On this basis, the internal force distribution of the wind barrier anchoring structure
under the most unfavorable train wind load is calculated, and the stress and structural damage stateof concrete near the U-bolt holes are evaluated.

Fig. 3. Finite element model: (a) Column of wind barrier and concrete pedestal; (b) U-shaped
anchor bolts.
3. Analysis of the Results
The judgment criterion of damage factor is introduced, as shown in Fig. 4. When the compressive
stress of concrete exceeds the peak value of compressive stress, the concrete is judged to becompression damage. When the tensile stress of concrete exceeds the peak value of tensile stress,
and the strength drops to 50% of the peak value, the concrete is judged to be tensile damage.
According to the specification and the simulated calibration test, the critical damage factors ofcompression and tension of C30 concrete are calculated, which are 0.325 and 0.624, respectively.

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1.0

1.0

0.8

0.8

Tensile damage factor

Compression damage factor

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

0.6
0.4

0.325

0.2
0.0
0.000

0.00147
0.002

0.004

0.006

0.008

0.010

Compression strain

0.012

0.6

0.5

0.4

0.2

0.0
0.0000

0.014

0.624

2.4E-40.00020.0004

Tension strain

0.00060.0008Fig. 4. Correspondence of damage factor and strain: (a) compression damage; (b) tensile damage.
Based on the structural dynamic response analysis, the displacement time-history curve of the windbarrier column is obtained, and it is found that the maximum displacement can reach 1.75mm andact on the top of the wind barrier column. The displacement time history was applied to thenonlinear finite element damage analysis of the structure, and the distribution nephogram of theconcrete tensile and compression damage factors at the bolt holes was obtained, as shown in Fig.
5. According to Fig. 5, under the combined action of train wind load and bolt pretightening force,the maximum tensile damage factor of surface concrete near the bolt hole is 0.833, while themaximum compressive damage factor is 0.022. The tensile damage factor exceeds 0.624,
indicating that the surface concrete is damaged by tension.

Fig. 5. Distribution nephogram of concrete tensile and compression damage factors at thelocationof U-bolt holes
4. Conclusions
Under the action of bolt pretightening force and train wind load, the anchor concrete near each bolt
hole has a risk of damage within a certain depth. Specifically, the peak value of tensile damage
factor of the surface concrete near the bolt hole on the windward direction was 0.833, and the peakvalue of compression damage factor was 0.022, and the surface concrete is damaged by tension at
this position. The concrete near the bolt hole has the risk of low cycle fatigue damage andanchorage damage under the action of train wind cycle load.
5. References
Xiong, X. H., Yang, B., Wang, K. W., Liu, T. H., He, Z., and Zhu, L. (2020). Full-scale experimentof transient aerodynamic pressures acting on a bridge noise barrier induced by the passage of highspeed trains operating at 380–420 km/h. Journal of Wind Engineering and Industrial
Aerodynamics, 204, 104298.
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Parametric Study on the Aerodynamic Characteristics
of Wind-Guide Barriers for Train-Bridge System
Shuo Jiang1*, Yunfeng Zou2, Xuhui He3, and Dianyi Guo4*
1: School of Civil Engineering, Central South University, Changsha and China; email:992334207@qq.com
2: School of Civil Engineering, Central South University, Changsha and China; email:yunfengzou@csu.edu.cn
3: School of Civil Engineering, Central South University, Changsha and China; email:xuhuihe@csu.edu.cn
4: School of Civil Engineering, Central South University, Changsha and China; email:guodianyi@csu.edu.cn*: corresponding author

Keywords: Wind-Guide Barriers; train-bridge system; aerodynamic characteristics; wind field;
wind tunnel test
Abstract: This paper presents a new type of wind guide barrier(WGB). Through wind tunnel
tests, the key parameters of three types of WGBs were studied, (including two flow guiding
parameters - the main wind deflector angle θ1 the sub-guide wind angle size h1, and one
shielding parameter - the root width B1 of the WGB), and their effects on the aerodynamic
characteristics of the high-speed train(HST) and blunt body box beam system and the bridge
deck wind field were investigated. To verify the windproof performance of the WGB, tests wereconducted with no barrier and a 30% ventilation rate rigid wind barrier. The aerodynamic forces
and wind pressure distribution of the train and bridge were obtained using force and pressure
measurement methods, and the wind field on the bridge was tested using a moving wind speed
testing system. The results show that the shielding effect is the main factor affecting thewindproof performance of the WGB. When the WGB is installed, the average wind pressure
distribution on the train is comparable to that of using a traditional wind barrier. At the sametime, the WGB can guide the oncoming flow, and the maximum ratio of the transverse windspeed and the vertical wind speed on the bridge deck can reach 0.25 and 0.17, respectively,
reducing the resistance of the WGB and the bridge-wind barrier system by 32% and 12%,
respectively. The change in the main wind deflector angle has a minor impact on the wind fieldon the bridge deck, and has no significant effect on the aerodynamic characteristics of the trainbridge system. The sub-guide wind angle size has a significant effect on the transverse andvertical wind guiding effects, reducing the sub-guide wind angle size increases the maximum
transverse wind speed ratio by more than 0.1, and also increases the maximum aerodynamic
force on the train. Increasing the root width of the WGB can enhance the shielding and guiding
effects, and can suppress the jet flow intensity near the bridge deck, which is beneficial for
windproof of the train.

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1. Introduction
The shielding effect of wind barriers creates a relatively safe local wind environment for HST(Kozmar et al., 2014). Researchers believe that the reduction of average wind speed and thevariation of turbulence are two important factors that affect the windproof performance (Takeda
et al., 1985; Dong et al., 2007, 2010). (Xiang et al., 2014) evaluated the wind-sheltering
performance of wind barriers using the DEA method, indicating that reasonable wind barrierheight and porosity would make HST have better aerodynamic performance, and other studies
have reached similar conclusions (He et al., 2014; Chu et al., 2013). However, we cannot only
focus on the windproof performance of wind barriers, and the study of bridge-barrier system isalso important. Wind barriers transmit wind loads to bridges, which significantly increase thebridge's drag coefficient, as well as the train's drag and lift coefficients on the bridge(Zhang et al.
2017). This paper studies a new type of WGB that deflects the airflow horizontally for
aerodynamic design. Through wind tunnel tests, the study investigates the effects of three WGB
parameters on the aerodynamic characteristics of HST and bridges. In addition, the wind fieldcharacteristics on the bridge deck were also tested, and the wind speed and turbulence field after
the wind barrier were analyzed, including the three-direction wind speed components andturbulence intensity, to comprehensively evaluate the windproof performance.
2. Outline of the Experimental Procedure
As shown in Fig. 1(a), 30 measuring points were arranged for each train section, and 42
measuring points were arranged for each bridge section. The wind speed point layout scheme isshown in Fig. 1(b), and a gradually denser layout is adopted from the top to the bottom. The
relevant parameters of the WGB have been listed in Table 1.
(a)

(b)

Fig. 1. Experimental details: (a) arrangements of pressure taps on the train and bridge crosssection; (b) arrangement of wind speed measuring points (Unit: mm)
3. Results and Discussion
Fig. 2 show the average and extreme aerodynamic forces of the train when it is located on thewindward sides of the bridge. The red and blue lines in Fig. 2 respectively represent the average
aerodynamic force of the train without barriers and with a traditional wind barrier installed.
When a type 4 WGB is installed, the average aerodynamic force of the train is close to that whena traditional wind barrier is installed, indicating that WGB can provide a similar windproof
performance for trains as traditional wind barriers. As shown in Fig. 2(a), when θ1 is slightly
adjusted, the average aerodynamic force of the train remains almost unchanged, but the extreme
aerodynamic force of the train is greatly affected by the angle of the windward surface. When
θ1=20°, the CH extreme value increases from 0.24 to 0.45, and at the same time, the CV extreme

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value increases from 0.52 to 0.71. As shown in Fig. 2(b), when B1 is increased from 0.25m to
0.3m, the average CH and average CV of the train will be reduced to varying degrees, but theextreme CH of the train will increase. At the same time, it should be noted that when B1 is further
Table 1. Types of WGB experiment
Width of WGB
DesignWind barrierrootparameters
types
of WGB
B1(mm)
B
1
h
2

1

θ2

H

θ1

B1

2

Angle of main
deflector plate
θ1(°)
No barrier

Size of the upper
deflector angle
h1(mm)

Rigid wind barrier (30% ventilation rate)

3

300

20

350

4

300

10

350

5

300

20

500

6

350

10

350

7

250

10

350

increased, the average aerodynamic force of the train does not change significantly, so the
optimal value of B1 should be between 0.25m and 0.3m. As shown in Fig. 2(c), although
increasing h1 can reduce the wind load on the WGB itself, both the average aerodynamic force
and extreme aerodynamic force of the train will increase. The maximum CV of the train increases
by 60%, which will increase the train's derailment coefficient. Overall, B1 and h1 are WGB
parameters that have a relatively large impact on the aerodynamic forces of the train, while θ1
has a relatively small impact. When optimizing the WGB parameters, the adjustment of the firsttwo parameters should be emphasized.
(c)Fig. 2. Effect of WGB parameters on HST aerodynamic force (upstream): (a) Angle of main
deflector plate θ1; (b) Width of WGB root B1; (c) Size of upper deflector angle h1
Fig. 3 illustrates the CV and CH of the bridge-barrier system, as well as the CH of the WGBs withdifferent installations. Generally speaking, the CH of the WGB roughly follows the law of
increasing with B1 and decreasing with θ1 and h1. Compared with the traditional vertical windbarrier, the installation of WGB (Type 4) with the same ventilation rate can reduce the maximum

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resistance and barrier resistance of the bridge-barrier system under various train-bridge
combinations by 12% and 32%, respectively.
The contour lines of the wind speed components in three directions under type 4 were given in
Fig. 4. From Fig. 4(a), it can be observed that the incoming flow is obstructed by the barrier andflows around, forming a shear layer above the barrier. On the other hand, due to the formation ofa jet through the gap at the bottom of the barrier, a high wind speed area with a wind speed ratioof more than 0.6 can be observed at the bottom of the barrier. From Fig. 4(b), it can be seen thatthe lateral wind speed basically shows a trend of decreasing as it moves further away from thebarrier. Affected by the jet, there is a high lateral wind speed near the bridge deck on thewindward side; At the top of the barrier and below the shear layer, there is a large range of lateral
airflow deflection, with a maximum lateral wind speed ratio of -0.25. The distribution of verticalwind speed is similar to that of crosswind wind speed, with an extreme value of 0.17 for thevertical wind speed ratio.
(c)Fig. 3. Aerodynamic forces of bridge and barrier: (a) Without train; (b) The train is locatedupstream; (c) The train is located downstream
ref

e

(a)

(b)

(c)

Fig.4. Mean wind speed contour lines of type 4: (a) streamwise wind speed; (b) transverse windspeed; (c) vertical wind speed. In all diagrams the wind blows from right to left-hand side.
4. Conclusion(1) The WGB provides comparable windproof performance for the train while reducing barrier
resistance and the resistance of the bridge-barrier system by 32% and 12%, respectively.
(2) The design parameters of B1 and h1 for the WGB have a significant impact on the average
aerodynamic forces of the train.
(3) The WGB can effectively reduce the wind speed on the bridge and has a strong guiding effecton the airflow. The maximum values of the transverse wind speed ratio and vertical wind speed

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ratio behind the WGB can reach 0.25 and 0.17, respectively.
5. References
He, X. H., Zou, Y. F., Wang, H. F., Han, Y., Shi, K. 2014. Aerodynamic characteristics of a
trailing rail vehicles on viaduct based on still wind tunnel experiments. Journal of Wind
Engineering and Industrial Aerodynamics, 135, 22–33.

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Earthquake and Dynamics

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Experimental Evaluation on Seismic Behavior of Precast Concrete Walls with
Developed Vertical Joints for Enhanced Shear Strength
Kyo Young Moon1, Sung Jig Kim2*, and Chunho Jang3
1: Department of Architectural Engineering, Keimyung University, Daegu, South Korea;
email: mky9404@gmail.com
2: Department of Architectural Engineering, Keimyung University, Daegu, South Korea;
email: sjkim4@kmu.ac.kr
3: Department of Civil Engineering, Keimyung University, Daegu, South Korea;
email: chunho@kmu.ac.kr*: corresponding author

Keywords: Precast concrete wall; Vertical joint; Cyclic loading test; Seismic performance
Abstract: This paper introduces the newly developed vertical joints between Precast Concrete (PC)wall panels to improve their in-plane shear capacity and ductility. The three types of vertical joints
are developed based on the existing vertical joints. The developed vertical joint systems are
responsible for increasing shear strength, stiffness, and shear transfer capacity between PC walls
through the dovetail-shaped metal laths, shear keys and reinforcing bars. To evaluate the seismic
performance of PC walls with developed vertical joints, four specimens including the conventionaland developed vertical connections were designed and constructed. The experimental results
showed that the PC walls with developed vertical joints not only had improved strength andductility, but also exhibited better crack and damage control, when compared to the specimen withthe conventional connection.
1. Introduction
The Precast Concrete (PC) structures has been widely used due to the advantages of high-quality
fabrication, construction cost savings, reduced material consumption, and rapid construction(Zhangfeng and Zhengxing, 2019; PCI, 2010). Due to limitations in transportation and lifting
equipment, there are generally many restrictions on the size of PC members. Therefore, in the caseof a structural wall having a large size compared to other members, the PC wall is usually
manufactured in a factory as a relatively smaller sized wall and thus, joints are inevitably createdbetween the PC walls during installation on size. The behavior of these joints can cause severe
damage to PC members under seismic loads and can have a significant impact on the overall
performance of the structure.
The mechanical behavior of the PC walls depends on the shear transfer capacity of the joints
between the precast components (Zhi et al. 2020). The connection details with inadequate force
transfer capacity and poor structural performance such as strength, ductility, and stiffness, could
cause severe damages and brittle failure of the entire structure when subjected to earthquake.
Therefore, in this study vertical connections between PC wall panels are newly developed toprevent brittle failure and provide sufficient in-plane shear and ductility. A total of four specimens

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were constructed and investigated through static cyclic loading tests. The seismic performance ofdeveloped connections is evaluated by comparing those of the wire loop connection.
2. Developed Vertical Connections of PC WallIn this study, three types of vertical connection were developed to improve upon the existing joint
by increasing the transverse reinforcement ratio and friction force at the joint interface, as
illustrated in Fig. 1. Typically, these connections are constructed by casting the mortar after
installing the reinforcing bars between the PC wall panels. The slit between the PC walls isdesigned to be 20mm according to Korea Design Standard (KDS 41 17 00).
The first connection is a dovetail-shaped metal lath with headed rebars as shown in Fig. 1(a). The
second and third connections are similar to the traditional connection as shown in Fig. 1(b), butwith a higher transverse rebar ratio of about 24.6%, as compared to the traditional connection (Leeet al. 2022). In the third connection, shear keys are added to the connection surface as shown Fig
1(b). The dovetail-shaped metal lath and shear keys are expected to enhance the stiffness and shear
strength of the vertical connection by increasing the friction forces on the connection surface. All
developed connections aim to improve the shear transfer capacity through the reinforcing bars.

(a)(b)
Fig. 1. Developed vertical connections; (a) Dovetail-shaped connection; (b) Connection withrebars and shear keys
3. Experimental Program
A static cycle loading test was conducted to evaluate the seismic performance of the PC walls with
developed vertical joints. Details of test specimens are summarized in the Table 1. All specimens
have the same geometric configuration except for the panel width, which varies depending on thetype of vertical connection. This is because the slits between the PC walls differ according to thetype of connection. The clear height and thickness of the PC walls are 2700 mm and 200 mm,
respectively. The specimen PWL, which has wire loop connections, was selected as the reference
structure for an existing vertical connection. The specimen PWD, PWR, and PWRS have theproposed connections in this study, and their descriptions are shown in Table 1. All specimens
were subjected to lateral cyclic loads with a constant axial load of 1,092kN, which represents 10%of the ultimate axial capacity of the wall (0.1𝑓𝑓𝑐𝑐′ 𝐴𝐴𝑔𝑔 ). The lateral displacement history was imposedwith multiple cycles at each displacement level to reflect the effect of strength degradation
characteristics.

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Table 1. Details of specimen
Specimen
Name

Geometric
configuration
(mm)
Panel
Slit
width
100
860

Description of connection

PWL

Wire loop
Dovetail-shaped metal lath and
PWD
20
900
reinforcing bars
PWR
Reinforcing bars
20
900
PWRS
Reinforcing bars and shear keys
20
900
𝑓𝑓𝑐𝑐′ , 𝑓𝑓𝑚𝑚 = concrete and mortar strength, respectively
𝜌𝜌𝑣𝑣 , 𝜌𝜌ℎ = vertical and horizontal reinforcement ratio, respectively

Material and section details
𝑓𝑓𝑐𝑐 ′
𝑓𝑓𝑚𝑚(MPa) (MPa)32.13 52.16

𝜌𝜌𝑣𝑣
(%)
0.33

𝜌𝜌ℎ
(%)
0.34

33.33

57.81

0.32

0.34

27.44
30.04

60.34
63.43

0.32
0.32

0.34
0.34

4. Experimental Results
Fig. 2 shows crack patterns of all specimens at the drift ratio of 1.0%. The average envelope curveand cumulative energy dissipation of all specimens are illustrated in Fig. 3. All specimens showeda similar crack pattern as shown in Fig. 2. The specimen PWL suffered the most severe damage
with large cracks. Among the specimens with developed vertical connections, the specimen PWR
had the most severe cracks and damage at the vertical connection. In contrast, in the case of thespecimen PWD and PWRS, few cracks were observed at the vertical connection.
Compared to the specimen PWL, the maximum lateral force of the specimen PWR decreased by
up to 2.42%, while those of specimens PWD and PWRS increased by 5.53% and 3.48%,
respectively. Furthermore, the cumulative energy dissipation capacity of the specimens with
developed vertical connection increased by up to 3.98%, when compared to the specimen PWL.

(a)
(b)
(c)
Fig. 2. Cracks patterns of specimens; (a) PWL; (b) PWD; (c) PWR; (d) PWRS

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(a)Fig. 3. Average envelope curve and cumulative energy dissipation(b)

5. Conclusion
In this study, newly developed vertical connections were proposed to improve the shear strengthof vertical connection and their seismic performance was evaluated through the static cyclic testing.
Compared to the specimen with a conventional connection, maximum strength and energy
dissipation capacity increased by up to 5.11% and 3.98%, respectively. In addition, the developed
connections were effective in damage and crack control. Based on the experimental investigation,
it could be concluded that the proposed vertical connections could improve the seismic
performance of PC walls.
6. Acknowledgments
This research was supported by a grant (RS-2021-KA161540) from Technology Business Innovation
Program funded by Ministry of Land, Infrastructure and Transport of Korean government.
7. References
Architectural Institute of Korea. 2019. In Seismic Building Design Code (KDS 41 17 00), Ministryof Land: Seoul, Korea
Dang, L., Liang, S., Zhu, X., Zhang, M., & Song, Y. 2021. Seismic performance of precast
concrete wall with vertical energy‐dissipating connection, The Structural Design of Tall andSpecial Buildings, 30(2), e1820.
Lee, W. J., Kim, M., Lee, D., Ju. H., & Kim, K.H. 2022. Seismic Performance Evaluation ofIntermediate Precast Structural Walls. Journal of Next-generation Convergence Technology
Association, 6(3), 421-432
PCI Industry Handbook Committee. 2010. PCI design handbook: precast and prestressed concrete.
In: Prestressed Concrete Inst, 7th ed., Chicago, IL
Zhangfeng, Z., & Zhengxing, G. 2019. Seismic performance of the spatial model of precast
concrete shear wall structure using grouted lap splice connection and cast‐in‐situ
concrete. Structural Concrete, 20(4), 1316-1327.
Zhi, Q., Xiong, X., Yang, W., Liu, S., & Xiong, J. 2020. Experimental study on the shear behaviorof precast wall concrete joints with/without dowel reinforcement. Materials, 13(7), 1726.

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Hybrid Testing of CFST Columns in Rigid-Frame Bridges Subjected to
Horizontal and Vertical Ground Motions
Riadh Al-Mahaidi1*, Javad Hashemi2, and Ali Al-Attraqchi3
1: Professor, Swinburne University of Technology, Melbourne, Australia,
email: ralmahaidi@swin.edu.au
2: Associate Professor, Swinburne University of Technology, Melbourne, Australia,
email: jhashemi@swin.edu.au3: former PhD student, Swinburne University of Technology, Melbourne, Australia,
email: ali.yasir.w@gmail.com*

corresponding author

Keywords: hybrid testing; CFST bridges; vertical ground motions
Abstract: In this paper, a hybrid simulation of a concrete-filled steel tube (CFST) rigid-frame
bridge structure is presented. The bridge structure is constructed using CFST columns of circularand square cross-section shapes, and it is subjected to combined horizontal and vertical groundmotions. The testing is performed using the multi-axis substructures testing (MAST) system at
Swinburne University of Technology in Melbourne, Australia. The results demonstrate theremarkable seismic performance of CFST columns, emphasizing their advantages in enhancing
the seismic resistance of critical and post-disaster bridges that are susceptible to strong verticalground motions.
1. Introduction
Assessing the seismic performance of bridge structures has been a difficult task because of theirsize and geometry. The conventional experimental methods such as shake table testing have
proven to be risky and impractical due to the possibility of structure collapse. A feasible alternative
approach is hybrid simulation, which involves experimental testing of the critical section of thestructure, while the remaining parts are simulated using numerical methods (Hashemi et al., 2016;
M. J. Hashemi and Al-Mahaidi, 2017; Hashemi et al., 2017; Del Carpio et al., 2017). Many studies
have used hybrid simulation for the seismic performance of bridge structures (Stathas et al., 2017;
Chae et al., 2017; Bousias et al., 2017), but most have focused on horizontal ground motions only,with limited data available on combined horizontal and vertical ground accelerations. The verticalcomponent can be significant at near-fault sites and could cause significant damage to reinforced
concrete (RC) columns. CFST members, which are known for their high ductility and energy
absorption capacity, have been proposed as a potential solution for piers susceptible to significant
vertical acceleration in bridge construction. Although CFSTs have been widely studied
experimentally, their behaviour under horizontal and vertical boundary forces is still not wellunderstood, making it challenging to anticipate their response during post-disaster operations. This
study aims to examine the seismic performance of circular and square CFST columns under
combined horizontal and vertical ground accelerations using hybrid simulation.

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2. Description of Rigid-Frame Bridge Structure
The bridge design was previously studied to investigate the effects of combined vertical andhorizontal ground motions on RC bridge piers using hybrid simulation tests by Kim et al. (2011).
The design was based on the western side of the United States in a seismic zone. The prototype
bridge structure was designed according to the Federal Highway Administration (FHWA) of the
U.S. Department of Transportation with two double-column bents and a total span of 97.6m
divided into three smaller spans, as shown in Fig. 1. The RC columns were replaced by CFST
columns with a clear height of 6m and the bridge deck consisted of a cast-in-place concrete box
girder with integral bent caps and laterally-restrained abutments. The columns were modelledusing beam-with-hinges elements in OpenSees and the plasticity model was calibrated using theresults of a 3-dimensional mixed-mode quasi-static cyclic experiment. Full details about the bridgeand modelling can be found in Al-Attraqchi et al. (2020).

Fig. 1. Numerical model of rigid-frame bridge, MAST system, and the circular and square columns
being tested
3. Hybrid Simulation
The Multi-Axis Substructure Testing (MAST) system in the Smart Structures Laboratory at
Swinburne provides a powerful tool for investigating the effects of extreme loading events on large
structural components using hybrid simulation testing (Hashemi et al., 2015). A sophisticated 6DOF control system utilizes eight high-capacity hydraulic actuators to enable the application ofcomplex multi-directional deformation or loading schemes to structural components.
During the hybrid simulation, the bridge structure was subjected to the Northridge 1994
Sylmar Converter station ground motion records, which have a high peak ground acceleration(PGA) of over 0.3g and a vertical-to-horizontal (V/H) ratio greater than 0.6. This ground motion
was responsible for significant damage to half of the bridges in the area and led to the collapse ordemolition of 9 bridges. The vertical component was scaled up to 0.748g to achieve a V/H ratio of
1.2, which has been seen in previous earthquake records. To cover the full response range, fiveintensity levels of combined horizontal and vertical ground motions were consecutively applied topush the structure to 0.25% (elastic), 2.0%, 4%, 6% and 8% maximum drift ratios, respectively.

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The hysteresis responses of circular and square columns are illustrated in Fig. 2 by plotting thebase shear versus the drift ratios; also, the axial forces of the columns are represented by the colour
map to emphasize the effect of the axial force level on the hysteresis response. It is clear, especiallyfrom the bigger cycles, that the 𝑃𝑃 − ∆ effects are significant, and the lateral force resistanceoscillates as a result of the axial force changing from compression to tension. Large hysteresis
loops are another indicator that suggests adequate ductility and energy dissipation without
approaching collapse. The collapse limit state is often defined as a percentage of loss from theultimate lateral strength, which was not observed in these columns.

Fig. 2. Hysteresis-response showing the shear force and lateral drift and the color showing axial
load
Other than a small residual displacement of about 50 mm, no major physical damage was found at
the end of the tests. The physical damage that was seen manifested as outward bulging at the top
and bottom of the columns in the region of the plastic hinge was brought on by theyielding/buckling of the steel and the crushing of the concrete. Fig. 3 displays the observed
damages at the top and bottom of both columns as well as the depth of the bulging. For the squareand circular columns, respectively, the greatest bulging depth was 7.2 mm, and it happened at the
top in the plastic-hinge zone for both columns. For the square column, the damage was outwardly
distributed to one side, whereas for the circular column, it was distributed to both sides.

Fig. 3. Observed damages at the end of hybrid simulation; (a) circular column top; (b) circularcolumn bottom; (c) square column top; (d) square column bottom242

Bridge Engineering Institute Conference 2023 (BEI-2023)
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4. Conclusion
This research investigated the seismic response of a three-span, double-column bridge constructedusing concrete-filled steel tube (CFST) columns, using a pioneering experimental technique called
multi-axis hybrid simulation testing. Two types of CFST columns, circular and square, were
examined and subjected to progressively increasing levels of Northridge ground motion. The
findings demonstrated that both types of columns exhibited excellent seismic performance, withductile behaviour, substantial hysteresis loops, and minimal physical damage. The circular column
performed slightly better than the square column. These findings underscore the potential of CFST
columns in constructing resilient bridges with minimal recovery costs and downtime, particularly
in near-fault locations subjected to extreme multi-directional seismic loads.
5. References
Al-Attraqchi AY, Hashemi MJ and Al-Mahaidi R (2020) Hybrid simulation of bridges constructedwith concrete-filled steel tube columns subjected to horizontal and vertical ground motions.
Bulletin of earthquake engineering 18(9): 4453-4480.
Bousias S, Sextos A, Kwon O-S, et al. (2017) Intercontinental hybrid simulation for the assessmentof a three-span R/C highway overpass. Journal of Earthquake Engineering. 1-22.
Chae Y, Park M, Kim C-Y, et al. (2017) Experimental study on the rate-dependency of reinforced
concrete structures using slow and real-time hybrid simulations. Engineering Structures 132:648-658.
Del Carpio RM, Hashemi MJ and Mosqueda G (2017) Evaluation of integration methods forhybrid simulation of complex structural systems through collapse. Earthquake Engineeringand
Engineering Vibration 16(4): 745-759.
Hashemi MJ, Al-Mahaidi R, Kalfat R, et al. (2015) Development and validation of multi-axis
substructure testing system for full-scale experiments. Australian Journal of StructuralEngineering 16(4): 302-315.
Hashemi MJ, Al-Ogaidi Y, Al-Mahaidi R, et al. (2017) Application of Hybrid Simulation for
Collapse Assessment of Post-Earthquake CFRP-Repaired RC Columns. Journal of StructuralEngineering 143(1).
Hashemi MJ, Mosqueda G, Lignos DG, et al. (2016) Assessment of Numerical and Experimental
Errors in Hybrid Simulation of Framed Structural Systems through Collapse. Journal of
Earthquake Engineering 20(6): 885-909.
Kim SJ, Holub CJ and Elnashai AS (2011) Experimental investigation of the behavior of RC
bridge piers subjected to horizontal and vertical earthquake motion. Engineering Structures33(7):2221-2235.
M. J. Hashemi HHTYA-OJLW and Al-Mahaidi R (2017) Collapse Assessment of Reinforced
Concrete Building Columns through Multi-Axis Hybrid Simulation. ACI Structural Journal114(2).
Stathas N, Skafida S, Bousias S, et al. (2017) Hybrid simulation of bridge pier uplifting. Bulletinof earthquake engineering 15(8): 3385-3398.
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Seismic Performance of Precast Segmental Bridge Columns with Resettable
Sliding Joints: Joint Responses and Simulation
Feng Liang1*, Yingqi Liu2, and Francis T.K. Au3
1: The University of Hong Kong, Hong Kong, China; email: liangf@connect.hku.hk: Wuhan University of Technology, Wuhan, China and The University of Hong Kong, Hong
Kong, China; email: selbyliu@whut.edu.cn
3: The University of Hong Kong, Hong Kong, China; email: francis.au@hku.hk
2*: corresponding author

Keywords: Concrete bridges; piers; precast segments; self-centering; simulation.
Abstract: Precast segmental bridge columns with planar sliding joints possess large
displacement capability and energy dissipation under quasi-static cyclic tests. Though the planar
sliding joints can increase the ability to survive in moderate to high seismicity regions, theadjacent column segments often result in excessive residual relative displacement after seismic
action, leading to costly and sometimes difficult repairs. The other concern is the joint resistanceagainst vehicle braking, collision and wind load on the bridge. To increase the self-centering
capability and lateral resistance of the planar joints, a system with 2-dimensional resettablesliding keys has been proposed. The innovative system adopts V- or W-shaped resetting guidekeys, a smooth segment interface with relatively low coefficient of friction, and partiallydebonded tendons that go through ducts installed inside the concrete component. This study
extends the conceptual design to a gentle 3-dimensional (3D) pyramid-shaped resettable joint
and conducts a numerical simulation of column joints subjected to lateral loading. The responses
of a resettable sliding joint include three Phases: (i) small flexural deformation before sliding
initiates; (ii) joint sliding; and (iii) activation of tendons locking upon reaching sliding capacity.
In addition, the effects of bond failure, material damage, and plasticity are also discussed. The
force-drift curves of the resettable sliding joint show: (a) the process of sliding initiation; and (b)the combined effects of bearing forces, friction resistance and concrete damage on the overall
stiffness. Comparison of the results shows that the proposed system has large initial stiffness,good energy dissipation and recentering capacity.
1. Introduction
Precast segmental bridge column (PSBC) is one of the Accelerated Bridge Construction
techniques aiming for fast and green construction and less traffic disruption. PSBC is classified
into emulative and non-emulative types. The non-emulative design can be further categorized as
pure rocking, hybrid rocking and hybrid sliding columns (Sideris et al. 2014) according to thejoint setting and energy dissipation methods (Zhang and Alam, 2020). In an attempt to promote
PSBC in moderate to high seismicity regions, resettable sliding joints (RSJ) with a partiallydebonded system have been proposed (Au et al. 2019, Liu et al. 2021). The system consists of Vor W-shaped resetting guide keys, which comprise interfaces with relatively low coefficient of244

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friction (CoF) to allow sliding, and ducts and duct adaptors that accommodate partially debonded
prestressing tendons, as shown in Fig. 2(a). This study extends the conceptual design to apyramidal-shaped joint with a convex component at the top interface and a concave part at thebottom interface in the 3D perspective (Fig. 1(a)). The duct adaptors have a larger diameter thanthe ducts, forming the unbonded region together with unrestrained tendons. With the 3D
configuration, the joint allows some relative joint sliding and some resetting tendency.
Resettable Joint

d
d s,i
β

Concave Shapecr

Duct Adaptor
Duct

N s,i

Resettable
Joint

(a)

εc

Nc

V R,f

(b)

Convex Shape

βrcr

Fig. 1. Design of RSJ. (a) 3D pyramid-shaped configuration; (b) Normal stress and strain
distribution at concrete surface and tendon forces.
(a)

N ext
V ext

(b)

M ext,tGrouting
Duct Adaptor

N ext

V ext

hda,total

(c)

V R,b

φ

vsl

V R,b
V R,l

usl

φl
usl,l

M ext,bFig. 2. Responses of RSJ subjected to external shear and moment. (a) Phase 1, before sliding
initiates; (b) Phase 2, sliding with both horizontal and vertical displacements; and (c) Phase 3,
sliding reaches the limit.
2. Joint Responses
Fig. 2 shows the forces acting on adjacent parts of RSJ in different phases. The axial force 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒
is assumed to be constant, while the shear force 𝑉𝑉𝑒𝑒𝑒𝑒𝑒𝑒 and moment 𝑀𝑀𝑒𝑒𝑒𝑒𝑒𝑒 are assumed to increase
monotonically. Before the initiation of sliding, i.e., Phase 1 in Fig. 2(a), the component of
resultant force along the slope is smaller than the static friction of the RSJ, leading to elastic
flexural deformation as a cantilever column. Both the axial force and moment are resisted by theconcrete section and the tendons. The corresponding resistant moment 𝑀𝑀𝑅𝑅 with respect to thecenterline of the joint as shown in Fig. 1(b) is:𝑑𝑑
1𝑑𝑑
𝑀𝑀𝑅𝑅 = 𝑀𝑀𝑐𝑐 + 𝑀𝑀𝑃𝑃𝑃𝑃 = � ×− 𝛽𝛽𝑟𝑟 𝑐𝑐𝑟𝑟 � 𝑁𝑁𝑐𝑐 + �(𝑑𝑑𝑠𝑠,𝑖𝑖 − ) 𝑁𝑁𝑠𝑠,𝑖𝑖
2 𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐
2
𝑖𝑖

(1)

where 𝑀𝑀𝑐𝑐 and 𝑀𝑀𝑃𝑃𝑃𝑃 are the resistant moment contribution from concrete and post-tensioned
strands, respectively; 𝛽𝛽 = inclination of RSJ; 𝑑𝑑 = depth of concrete section; 𝑐𝑐𝑟𝑟 = contact depth;𝛽𝛽𝑟𝑟 = stress distribution factor determining the location of concrete contact force 𝑁𝑁𝑐𝑐 ; 𝑑𝑑𝑠𝑠,𝑖𝑖 =245

Bridge Engineering Institute Conference 2023 (BEI-2023)
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distance between tendon No. i and extreme compression fiber of segment and 𝑁𝑁𝑠𝑠,𝑖𝑖 = thecorresponding tendon force. The external axial force 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒 and the total prestressing force 𝑁𝑁𝑃𝑃𝑃𝑃 areresisted by the concrete, leading to:
𝑁𝑁𝑐𝑐 𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐 + 𝑉𝑉𝑅𝑅,𝑓𝑓 𝑠𝑠𝑠𝑠𝑠𝑠𝑠𝑠 = 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒 + 𝑁𝑁𝑃𝑃𝑃𝑃 = 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒 + � 𝑁𝑁𝑠𝑠,𝑖𝑖
𝑖𝑖(2)

where 𝑉𝑉𝑅𝑅,𝑓𝑓 is the friction force, a portion of the total shear resistance force 𝑉𝑉𝑅𝑅 of RSJ. In Phase 1,
the external axial force 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒 is constant, while the prestressing force 𝑁𝑁𝑃𝑃𝑃𝑃 and the corresponding
moment 𝑀𝑀𝑃𝑃𝑃𝑃 also remain the same as the initial values 𝑁𝑁𝑃𝑃𝑃𝑃,𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖 and 𝑀𝑀𝑃𝑃𝑃𝑃,𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖 , respectively.

Once the static friction force is overcome, the RSJ starts relative motion with both horizontal andvertical displacement 𝑢𝑢𝑠𝑠𝑠𝑠 and 𝑣𝑣𝑠𝑠𝑠𝑠 , respectively, due to the slope. The process is shown as Phase 2
in Fig. 2(b), where 𝑣𝑣𝑠𝑠𝑠𝑠 = |𝑢𝑢𝑠𝑠𝑠𝑠 |𝑡𝑡𝑡𝑡𝑡𝑡𝛽𝛽. Assuming negligible deformation of tendons in the bonded
region, the sliding leads to a deviation angle 𝜑𝜑 between the unbonded region of tendon and thevertical as well as vertical elongation of tendon 𝑣𝑣𝑠𝑠𝑠𝑠 . The bearing forces between the unbonded
tendons and the sides, 𝑉𝑉𝑅𝑅,𝑏𝑏 , along the slope also contribute to the shear resistance 𝑉𝑉𝑅𝑅 . The
deviation angle 𝜑𝜑 can be estimated from:𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡 =|𝑢𝑢𝑠𝑠𝑠𝑠 |𝑠𝑠𝑠𝑠𝑠𝑠(𝑢𝑢𝑠𝑠𝑠𝑠 )ℎ𝑑𝑑𝑑𝑑,𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡 + 𝑣𝑣𝑠𝑠𝑠𝑠(3)

where ℎ𝑑𝑑𝑑𝑑,𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡𝑡 = total height of two adjacent duct adaptors. Owing to the elongation upon1

initiation of sliding, the updated tendon force is 𝑁𝑁𝑃𝑃𝑃𝑃,𝑠𝑠𝑠𝑠 = 𝐸𝐸𝐴𝐴𝑃𝑃𝑃𝑃 �𝑐𝑐𝑐𝑐𝑐𝑐𝜑𝜑 − 1� + 𝑁𝑁𝑃𝑃𝑃𝑃,𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖𝑖 , where 𝐸𝐸
and 𝐴𝐴𝑃𝑃𝑃𝑃 are the Young’s modulus and cross-sectional area of tendons, respectively.

As sliding reaches the limit, 𝑢𝑢𝑠𝑠𝑠𝑠,𝑙𝑙 , in Phase 3 with the unbonded tendons getting into full contact
with the duct adaptors, as shown in Fig. 2(c), the joint is subjected to the force induced by dowel
action, 𝑉𝑉𝑅𝑅,𝑙𝑙 . The sliding limit 𝑢𝑢𝑠𝑠𝑠𝑠,𝑙𝑙 is determined by the radii of duct adaptor and tendon, i.e.,
𝐷𝐷 −𝐷𝐷𝑢𝑢𝑠𝑠𝑠𝑠,𝑙𝑙 = 𝑑𝑑𝑑𝑑 2 𝑃𝑃𝑃𝑃. The total shear resistance 𝑉𝑉𝑅𝑅 along the slope is:
𝑉𝑉𝑅𝑅 = 𝑉𝑉𝑅𝑅,𝑓𝑓 + 𝑉𝑉𝑅𝑅,𝑏𝑏 + 𝑉𝑉𝑅𝑅,𝑙𝑙 = 𝜇𝜇𝑁𝑁𝑐𝑐 𝑠𝑠𝑠𝑠𝑠𝑠(𝑢𝑢̇ 𝑠𝑠𝑠𝑠 ) + 𝑁𝑁𝑃𝑃𝑃𝑃,𝑠𝑠𝑠𝑠 sin(𝜑𝜑 + 𝛽𝛽) +𝐾𝐾𝑙𝑙〈|𝑢𝑢 | − 𝑢𝑢𝑠𝑠𝑠𝑠,𝑙𝑙 〉𝑠𝑠𝑠𝑠𝑠𝑠(𝑢𝑢𝑠𝑠𝑠𝑠 ) (4)𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐 𝑠𝑠𝑠𝑠

where 𝐾𝐾𝑙𝑙 is the horizontal stiffness due to dowel action; 〈𝑢𝑢〉 is the ramp function defined as〈𝑢𝑢〉 = 𝑢𝑢 for 𝑢𝑢 ≥ 0, and 〈𝑢𝑢〉 = 0 for 𝑢𝑢 < 0, giving the dowel force upon closure of the gap. The
total resisting shear force in Phase 3 comprises three components, i.e., friction force 𝑉𝑉𝑅𝑅,𝑓𝑓 , bearing
force 𝑉𝑉𝑅𝑅,𝑏𝑏 and dowel action 𝑉𝑉𝑅𝑅,𝑙𝑙 . On the one hand, the inclination 𝛽𝛽 contributes a certain amountto the shear resistance according to Eq. (4). On the other hand, the external forces along the slope
is 𝑉𝑉𝑒𝑒𝑒𝑒𝑒𝑒 𝑐𝑐𝑐𝑐𝑐𝑐𝜑𝜑 − 𝑁𝑁𝑒𝑒𝑒𝑒𝑒𝑒 𝑠𝑠𝑠𝑠𝑠𝑠𝜑𝜑. Therefore, part of the external axial force also acts as the restoring
force and thus increases the stiffness of the system.
Eqs. (1)-(4) will degenerate to the case of flat interface by setting 𝛽𝛽 = 0 as described by Sideris
et al. 2014. The three phases are simulated ignoring any bond failure of the partially debonded
system. The bond strength governs when bond slip takes place, considering that the grout is

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subjected to bearing stresses, dowel action and the increasing tendon force. The stiffness of RSJ
will decrease significantly after bond failure. A numerical study is conducted using ABAQUSCAE to evaluate the performance of resettable sliding joint in the three phases and the influenceof bond failure. Two numerical models are examined here, i.e., the RSJ with a 6° inclination andthe reference model with flat interface.

50
0
-50

Force = 0

-100
(a)

-30

Bond Slip

100

-20

Top Displacement (mm)
-10
0
10
20

0 deg
6 deg
30

Reaction Force (kN)

100

Reaction Force (kN)

3. Simulation
The 3D RSJ model, as shown in Fig. 1(a), comprises two concrete segments of size 0.35 m ×
0.35 m × 0.4 m with four prestressing tendons each having a cross-sectional area of 140 mm2,
where the bonded regions are modelled as cohesive contact. Embedded constraints are adopted at
the two ends of tendon, which simulate the anchorages and ensure that the tendon keeps working
even after bond failure. The bottom face of the bottom segment of RSJ is in dry contact with anelastic concrete base having a high CoF of 0.9, which effectively allows minor rocking but nosliding. The RSJ interface has a low CoF of 0.1 to facilitate sliding. The unbonded parts oftendons may be under bearing stresses and dowel action. A concrete damaged plasticity modeland a plastic strand model are adopted to evaluate the performance. The top face of the joint issubjected to cyclic lateral loading under displacement-control to five levels ranging from 0.44%
to 3.5% drift ratio based on the total height of 0.8 m, while the external axial force is 100 kN.

50
Dowel Action
10(b)

Bond Slip20
usl,l =17
Top Displacement (mm)Fig. 3. Hysteretic response of 0° sliding joint and 6° resettable sliding joint. (a) Lateral reaction
force versus top displacement; (b) Enlarged part of Fig. 3(a) showing bond slip and dowel action.
Fig. 3(a) shows the responses of a 6° RSJ and a 0° flat sliding joint (SJ). In Phase 1, there is no
relative sliding for the two joints and thus they have the same stiffness. However, as described in
Eq. (4), the RSJ has higher shear resistance than that of the flat SJ. The figure shows that thelateral force to initiate sliding of the RSJ is nearly twice that of the flat SJ. For the top drift
between 3.5 to 15 mm in Phase 2, the stiffness of the RSJ is also larger than that of the flat SJ. At
the initial stage of Phase 2 for the flat SJ, only friction resists the external force and the stiffness
rises with the increasing deviation of tendons due to sliding as shown in Eq. (3). However, nohardening phenomenon is observed for the RSJ in Phase 2, because with the tendon elongationand hence the rising tendon force, the resulting concrete damage at the two embedded ends
somewhat balances the hardening effects of tendons. The flat SJ enters Phase 3 when sliding
reaches 24 mm as shown in Fig. 3(b), when the dowel force significantly increases the shear
resistance. Phase 3 for the flat SJ appears to be delayed when the top displacement exceeds thesliding limit as rocking at the bottom joint also contributes to the top drift. However, before the
RSJ enters Phase 3 as shown in Fig. 3(b), bond failure occurs due to the rising tendon force,

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causing severe concrete damage. Therefore, the stiffness of the RSJ decreases without any
hardening phenomenon due to dowel action. Fig. 3(a) shows that the RSJ has much better
recentering capability upon unloading despite the concrete damage and plastic tendon
deformation in the final loading cycle, while the flat SJ has much bigger residual displacement.
The drift-force curves for the RSJ in Fig. 3(a) also enclose relatively large areas, suggesting goodenergy dissipation of the system.
In summary, this numerical study has obtained the three phases of responses of resettable and flat
sliding joints to imposed cyclic lateral displacement. With the V-shaped resetting guide keys, theRSJ performs better in respect of initial stiffness, energy dissipation and resetting capability. The
first aspect is essential for serviceability when relative joint displacement is undesirable undervehicle braking, collision and wind load on bridges. The excellent energy dissipation and selfcentering capability of the RSJ during an earthquake will help not only to ensure structural safety
but also to reduce repair costs.
4. References
Au, F.T.K., Liu Y.Q., Yuan, Y., Su, R.K.L., and Lam, N.T.K. 2019. Seismic performance of
precast segmental bridge columns with resettable sliding joints: feasibility study, Proceedings ofthe Bridge Engineering Institute Conference, 594-598.
Liu, Y.Q., Yuan Y., Liang, F. and Au, F.T.K. 2021. Resetting response of precast segments withgently keyed interface under seismic action, Earthquake Engineering & Structural Dynamics, 50(2), 601-618
Sideris, P., Aref, A.J. and Filiatrault, A. 2014. Quasi-static cyclic testing of a large-scale hybrid
sliding-rocking segmental column with slip-dominant joints. Journal of Bridge Engineering 19(10), 04014036
Zhang, Q. and Alam, M.S. 2020. State-of-the-Art Review of Seismic Resistant Precast Bridge
Columns. Journal of Bridge Engineering, 25(10), 03120001

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Effects of Vertical Ground Motions on
the Response of Long-Span Prestressed Bridges
Hsiao-Hui Hung1*, Chang-Wei Huang 2, and Yi-Hui Lu2
1: National Center for Research on Earthquake Engineering, Taipei, Taiwan; email:hhung@narlabs.org.tw
2: Department of Civil Engineering, Chung Yuan Christian University, Taoyuan, Taiwan*: corresponding author

Keywords: near-fault earthquakes; time history analysis; balanced cantilever bridge
Abstract: The purpose of this study is to investigates the vertical seismic effects on the longspan prestressed bridge, and further to explore the conservatism of domestic seismic designcodes for bridge in Taiwan for vertical earthquakes. The influence of vertical seismic effect on
the seismic performance of a representative bridge is discussed by numerical analysis method.
From the analysis, the differences between the results obtained by different analysis methods
suggested in domestic code are compared, and the results are also compared with those obtainedby adopting actual near-fault earthquakes as input through time history analysis. The analysis
results show that the bridge internal force due to vertical ground excitations obtained by thestatic analysis method in the domestic bridge seismic design code is more conservative than thatobtained by the dynamic method. In addition, the vertical seismic effects of near-fault
earthquakes are strongly dependent on the frequency content of the earthquake in the verticaldirections.
1. Introduction
In the early seismic design of bridges, vertical ground motions were mostly not taken into
account. However, in recent years, as countries around the world have successively sufferedfrom severe structural damage due to near-fault earthquakes, people noted that in addition to alarger velocity pulse, another notable difference between near-fault earthquakes and far-fieldones is that the near-fault earthquakes are often accompanied by larger vertical excitations.
Therefore, the vertical seismic effect has gradually attracted people's attention (Button et al.
2002) and has also been included in the seismic design of the structures recently. However, there
is still no consistent standard for the consideration of vertical seismic effects in the seismic
design codes of bridges in various countries. Until a more complete study is available, theTaiwan seismic design code for bridges revised the vertical design seismic forces withconservative considerations in 2019.
According to the Taiwan’s seismic design provisions for highway bridges (MOTC 2019),
for bridges located in near-fault zone, the effect of vertical seismic force should be considered. In
this design code, vertical earthquake adopts the same design response spectrum as horizontal
earthquake, but with a smaller magnitude. The magnitude of vertical earthquake is assumed to be
2/3 times the horizontal one. The seismic analysis methods can be categorized into static analysis

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method and dynamic analysis method. For static analysis method, the lateral design force fordesign earthquake is the value of the design response spectrum at the fundamental period of theconsidered bridge multiplied by its gravitational load. The vertical design seismic force can bedivided into contribution from superstructure and substructure. For the superstructure, theseismic force is the weight of superstructure multiplied by the plateau segment values of thevertical seismic design response spectrum. For the substructure, the seismic force is the weight ofthe substructure multiplied by the vertical spectral acceleration at period of zero. The dynamic
analysis method includes elastic response spectrum method, elastic time history analysis method
and nonlinear time history analysis method. For response spectrum method, the input response
spectra for both horizontal directions and vertical direction are the same as that for the static
analysis. For time history analysis method, all input ground motions shall be matched to thedesign spectrum as mentioned above using the time domain method.
In summary, it can be found that Taiwan’s seismic design code adopts the same response
spectrum for vertical ground motions and horizontal ground motions, but the magnitude ofvertical one is taken as 2/3 of that of the horizontal one. In addition, for the static analysis,
regardless of the fundamental vertical period of the bridge, the value of the design spectral
acceleration conservatively adopts the value of the plateau segment of the design response
spectrum, which is also the maximum value of the spectrum. In order to explore theconservatism of domestic bridge seismic design codes for vertical earthquake, and further tounderstand the actual impact of vertical ground excitations on the seismic performance ofbridges, this study takes a long-span RC prestressed bridge as an example to numerically study
this issue.
150m

80m

M

80m

R

P1

M

R

P4

31.92mP2

P3

Fig. 1. Layout of the considered bridge
2. Description of the Considered Bridge
The representative bridge considered in this study is a three-span segmental, cast-in-place
concrete cantilever bridges with a span configuration of 80m+150m+80m, as shown Fig. 1. The
substructures are all hollow elliptical RC columns. Among them, Piers P2 and P3 are integral
with the superstructure, and Piers P1 and P4 are connected to the superstructure by movable
constraint in the longitudinal direction and hinge constraint in the transverse direction. To
conduct the seismic analysis, a structural model of the bridge was established by commercial
software Midas/Civil. This model includes 69 construction segments and 258 continuousprestress tendons. In addition, time-dependent behavior of concrete and tendon was consideredusing construction stage analysis. Using this analysis model, the modal analysis was conducted
at first and the fundamental structural periods for the bridge were obtained. The obtained
fundamental periods in the longitudinal (x), transverse (y) and vertical (z) direction are 1.46sec,
1.55sec and 1.1sec, respectively.

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3. Seismic Analysis
The scope of the study involves linear analyses of the representative bridge by static analysis
method, response spectrum method and linear time history analysis method according to the
local design code. The analysis results are then compared with time history analysis under the
input of two actual near field ground motions to investigate the conservatism of domestic bridge
seismic design codes for vertical earthquakes.
The considered bridge is assumed to be located in Hualien City, Hualien County, whichis among the most seismically active areas in Taiwan. Fig. 2 shows the design response spectraof Hualien City, which will be used in the static analysis and response spectrum analysis in thisstudy. Also given in Fig.2 is the matched spectrum of the compatible earthquake matched to thesite’s design spectrum code. In order to figure out the difference between the behavior of thebridge under the actual near-fault earthquake and the analysis results according to the code, thisstudy selected two real earthquake records for dynamic time history analysis. One was therecords recorded at station HWA009 during 2018 Hualien earthquake (Richter magnitude scale=6.3), and the other was recorded at station TCU068 during 921 earthquake (Richter magnitude
scale =7.3). The response spectra for these two real earthquake records are also shown in Fig.2.

Acceleration (gal)

2000
1500

matched spectrum
design spectrum
HWA009TCU068Acceleration (gal)

Z (+)

2500

1000
500
0

0

0.5

1

1.5
2
Period (Sec)

(a)

2.5

3

matched spectrum
design spectrum
HWA009TCU068

N (+)

1200
1000
800

Acceleration (gal)

1400

3000

600
400
200
0

0

0.5

1

1.5
2
Period (sec)

(b)

2.5

3

1800
1600
1400
1200
1000
800
600
400
200
0

matched spectrum
design spectrum
HWA009TCU068

E (+)

0

0.5

1

1.5
2
Period (Sec)

2.5

3

(c)

Fig. 2. Response spectrum in (a) vertical direction; (b) north-south direction; (c) east-west
direction.
4. Analysis Results
Fig. 3 shows the member forces of bridge superstructure obtained by different analysis methods
due to seismic forces. In these figures, H represents the analysis results considering only
horizontal ground excitations, and H+V represents the analysis results considering bothhorizontal and vertical ground excitations. To account for the directional uncertainty ofearthquake motions, a combination of seismic forces following 30% combination rule wasadopted for the static analysis and response spectrum analysis. For the time history analysis,
simultaneous occurrence of ground motions in different directions were considered, including
simultaneous input of two horizontal ground accelerations (H case), or simultaneous input of two
horizontal and one vertical ground accelerations (H+V case). By comparing the results of cases
H and H+V, we can find that the influence of vertical earthquakes on the superstructure member
forces of such a three-span concrete cantilever bridge is mainly in the bending moment My andshear force Vz at the location close to the middle supports, and the axial forces at the middle span.
Vertical earthquake has no effect on the other internal forces of the superstructure, including the
moment Mz, shear Vy along the transverse direction and torque, so they are not shown in thispaper. In the comparison of results obtained by different analysis methods, the static analysis

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method as expected is much more conservative than the response spectrum analysis method. In
terms of the time history analysis results of real near-fault earthquakes, the vertical seismic effectof HWA009 is quite insignificant, but on the contrary, the vertical earthquake effect of TCU068
is relatively much larger. The observation mainly due to the fact that the response spectral valueof HWA009 at the period of 1.1 sec, that is the vertical fundamental structural period of theconsidered bridges, is very small. On the other hand, the response spectral value of TCU068 at
the period of 1.1sec is much larger. In the comparison with the real near-fault earthquake action,
the results of the three analyses according to the code are greater than the results under the actionof HWA009, but the results under the action of TCU068 are underestimated. This result indicatesthat the effect of near-fault earthquake on the seismic performance of the long span PC bridge ishighly dependent on the site-specific characteristic of the records. Therefore, more research isneeded before complete conclusions can be drawn.

200

300

-100000

Position (m)

0

-1000

200

300

0

200

Position (m)

300

-3000

100

H
H+V

1500

Position (m)

1000
500
0
-500
-1000
-1500

300

1500
500
0
-500
-1000

0

100

200

Position (m)

300

100

200

Position (m)

300

200

Position (m)

300

-3000

H
H+V

1000
500
0
-500

Position (m)

300

H
H+V

1000
0

-3000

100

200

Position (m)

300

0

100

1000
500
0
-500

200

Position (m)

3500

H
H+V

1500

-2500

300

-2000

0

300

H
H+V

2500
1500
500
-500
-1500
-2500

-2000
200

200

-1000

-1500

100

100

Position (m)

2000

-1000

0

0

3000

0

2000

1500

-2500

-100000

H
H+V

2500

-2000

100

Position (m)

300

-2000

0

-1500

0

200

-1000

-1000

-1500
-2500

-3000

-80000

100

1000

0

0

-60000

0

2000

2000

1000

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-2000

H
H+V

2500

H
H+V

2000

Axial force (tonf)

2000

200

-100000

20000-40000

3000

-2000

0

2500

2500

300

-1000

Axial force (tonf)

100

Position (m)

200

1000

-2000
0

100

2000

1000

-1000

-2000

0

40000

-20000-80000-100000

3000

H
H+V

2000

1000

Vz (tonf)

Vz (tonf)

100

0-60000-80000020000-20000-40000-60000

3000

H
H+V

2000

Axial force (tonf)-40000Vz (tonf)

100

3000

-2500

0

-800000Position (m)

-3000

20000
-20000My(tonf-m)-40000-60000-80000-100000

0

H
H+V

60000

40000

Vz (tonf)-6000020000-20000

60000

80000

Axial force (tonf)-4000040000

Time history analysis (TCU068)

100000

H
H+V

80000

Vz (tonf)

0-20000

60000

40000

Time history analysis (HWA009)

100000

H
H+V

80000

Axial force (tonf)

20000

60000

My(tonf-m)

40000

Time history analysis (code compatible EQ)
100000

H
H+V

80000

My(tonf-m)

My(tonf-m)

60000

Response spectrum method

100000

H
H+V

80000

My(tonf-m)

Static analysis method
100000

0

100

200

Position (m)

300

-3500

0

100

200

Position (m)300

Fig. 3. Comparison diagram of the influence of vertical earthquake on member forces obtainedby different analysis methods.
It should be noted that the above analysis results are based on linear analysis, that is, without
considering the nonlinearity of the pier, if the nonlinearity of the pier is taken into account, theinfluence of the vertical seismic force will be more prominent due to the reduction of the internal
forces of superstructure that are induced by lateral forces after the connected piers yielded.
5. Conclusions
This study investigates the vertical seismic effect on the seismic performance of a long spanprestress concrete bridge using numerical analysis method. The analysis results show that the
major influence of vertical seismic forces is on the bending moment of the main girder at theintermediate supports. In addition, the vertical seismic effects of near-fault earthquakes arestrongly dependent on the site-specific characteristic of the earthquake.

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6. References
Button, M.R., Cronin, C.J., and Mayes, R.L., 2002, Effect of vertical motions on seismic
response of highway bridges, Journal of Structural Engineering, 128(12), 1551-1564.
MOTC, 2019, Seismic Design Code for Highway Bridges. (in Chinese).

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Seismic Performance Assessment of Wall Piers Supported by Pile Foundation
Shin-Tai Song1* and Wen-Li Huang2
1: Associate Professor, Department of Civil Engineering, National Chung-Hsing University,
Taichung, Taiwan, R.O.C.; email: ssong@nchu.edu.tw
2: Graduate Research Assistant, Department of Civil Engineering, National Chung-Hsing
University, Taichung, Taiwan, R.O.C.; email: wendy310322@gmail.com*: corresponding author

Keywords: seismic performance; pile foundation; wall pier; capacity spectrum approach
Abstract: Piles in foundation are designated as the critical members of wall pier bents undertransverse seismic loads. In this paper, the seismic performance of a wall pier bent is assessed using
a nonlinear static procedure based on capacity spectrum approach. The approach uses nonlinear
pushover analysis to assess the capacity of a wall pier and design spectrum to assess seismic demand.
The ductility capacity of the wall pier at its performance limit state is compared with the ductility
demand imposed by the design level seismic load. Satisfactory seismic performance can be
guaranteed if the ductility demand does not exceed the ductility capacity. The effect of riverbed
scouring on seismic performance of wall pier is also illustrated in this paper.
1. Introduction
Wall piers are widely used as the substructure for bridges with a relatively narrow deck and tall piers.
Wall piers provide high lateral stiffness to ensure stability in the transverse direction of bridges. Given
that this type of bridge pier behaves like a rigid wall under lateral loads, the primary damage
mechanism is brittle shear failure, which must be avoided. A newly issued seismic performance
evaluation specifications for highway bridges in Taiwan (MOTC 2020) designates the piles in thefoundation as the critical members of wall pier bents under transverse seismic loads. A limited levelof inelastic deformation is permitted at the pile/pile–cap interface. Brittle shear failure is prevented byensuring that the shear strength of the wall pier is sufficiently larger than the lateral strength of thefoundation. The seismic performance of a wall pier supported by a pile foundation can be assessed bycomparing its ductility capacity with the ductility demand under the design level earthquake loads.
2. Method for Seismic Performance Assessment
In this study, the performance of wall piers under transverse seismic loads is assessed using ananalytical procedure based on capacity spectrum approach, which was proposed in ATC-40 (ATC1998) and later enhanced in FEMA-440 (NEHRP 2005). The capacity spectrum of the bridge bent
is constructed based on the lateral pushover curve through an approach recommended in ATC-40(ATC 1998). The points corresponding to the performance limits of the bridge are identified in thecapacity spectrum to ensure satisfactory performance of the bridge bent. The ductility capacity ofthe bridge bent is equal to the ratio of the displacements at the specified performance limit and theeffective yield limit state. The seismic demand on the structure is assessed using the acceleration–
displacement design response spectrum, which is constructed based on the seismic demand in the254

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specification (MOTC 2020). The damping effect from structural yielding and the associated seismic
demand reduction must be considered during the construction of the demand spectrum. The ductility
demand imposed on the bridge bent is determined by the point where the demand spectrum intersects
with the capacity spectrum. The detailed procedure for finding the displacement ductility demandof the structure is available in literature (ATC 1999; FEMA 2005). The seismic performance can beassessed by comparing the ductility demand with the ductility capacity at the performance limit.
3. Wall Pier Bent and Finite Element Model
A multi-span bridge in the West Coast Expressway in Taiwan is selected as the illustrative example.
The geometry of the wall pier and the pile foundation is plotted in Fig. 1. The compressive strengthof the concrete is f 'ce = 45 MPa . The yield strength of reinforcements is f y = 475 MPa . The finite
element model of the bridge bent is constructed in the earthquake engineering simulation platform
OpenSees and shown in Fig. 1. The wall pier is modeled by elastic elements with extremely highstiffness. The piles are modeled by beam–column elements with fiber section. Because the piles aredesignated as the seismic critical members for wall pier bent, the performance limit is defined as theinelastic deformation at the pile/pile–cap interface reaching the damage-control limit state. The
strains recommended for the damage-control limit state to avoid subsequent replacement of thestructure are available in ACI318-19 (ACI 2022), with a confined concrete compressive strain limitof 0.015 and a longitudinal steel tensile strain limit of 0.050. The bridge is assumed to be located at
a medium sand site, with an effective friction angle of φ = 35o . In the finite element model, the soil
resistance on piles is modeled by three types of spring, representing the friction and lateral forces
applied on the pile and bearing resistance at the pile tip. The bridge is originally designed with thepiles fully embedded in the soil, that is, La = 0 m . However, the bridge suffers from the pile
exposure of La = 6 m caused by riverbed scouring.

Fig. 1. Geometry and finite element model of a bridge bent
4. Capacity and Demand Spectra
The capacity spectrum of the wall pier bent is constructed based on the load–displacement curvefrom monotonic pushover analysis. Figure 2 shows the capacity spectra of the wall pier bent before
riverbed scouring (i.e., La = 0 m ) at a scour depth of La = 6 m . The equivalent yield limit state of255

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the wall pier is determined by the bilinear idealization of the capacity spectrum. The ductility
capacity of the wall pier bent at the damage-control limit state is µcap =1.89 when the piles are
fully embedded. The overturn of the wall pier under lateral loading imposes large axial force on theleading row of the piles. The figure shows that the piles in the first row of the foundation are only
subjected to an axial force of P1st row = 3185 kN before the lateral push load is applied. The axial
force increases about 6 times to P1st row = 18895 kN when the pile is at the damage-control limit state.
At the scour depth of La = 6 m , the ductility capacity of the wall pier bent at the damage-control
limit state reduces to µcap =1.75 . The axial force on the leading row of the pile is P1st row = 12537 kN .
In view of the effect of soil nonlinearity, the effective damping ratio at yield limit state can be
determined by performing cyclic pushover analysis of the wall pier bent with a lateral displacement
up to the equivalent yield displacement. The effective damping ratios at the yield limit state are
estimated to be ζ el =10% for La = 0 m and ζ el =8.5% for La = 6 m . The 5% damping
acceleration–displacement design response spectrum correlated to the maximum considered
earthquake (MCER) at the bridge site is plotted in Fig. 2. The response spectra at damping ratios ofζ =8.5% and ζ =10% are also plotted in Fig. 2 and used to construct the demand spectrum forseismic performance assessment.

Fig. 2. Capacity spectra of the bridge bent and MCER acceleration–displacement spectrum of the site
5. Seismic Performance Assessment of the Wall Pier Bent
The seismic performance of the wall pier bent is assessed by capacity spectrum approach. An
iterative process is recommended in ATC-40 (1999) for finding the point where the demand
spectrum intersects with the capacity spectrum. The demand spectrum needs to be scaled downbecause of the damping effect caused by the inelastic deformation of the wall pier bent. Specifically,
FEMA 440 (NEHRP 2005) suggests a correlation between the inelastic deformation, characterized
by the displacement ductility factor µ , and hysteretic damping ζ h . The equivalent damping ζ eq of
the yielded wall pier bent is the sum of effective damping ζ el at the yield limit state and hysteretic
damping ζ h after the yield limit state. The damping coefficient BD from ASCE7-16 (ASCE 2017)256

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is used for scaling the demand spectrum. The results of the seismic performance assessment areplotted in Fig. 3. For the scour depth of La = 0 m , the demand spectrum intersects the capacity
spectrum at a displacement ductility demand of µ dem =1.10 , which provides a hysteretic dampingof ζ h =0.05% . The equivalent damping of the wall pier bent is ζ eq =10.05% , and the damping
coefficient is BD = 1.201 . For the scour depth of La = 6 m , the displacement ductility demand isassessed to be µ dem =1.58 , providing a hysteretic damping of ζ h =1.43% . The equivalent dampingof the wall pier bent is ζ eq =9.93% , and the damping coefficient is BD = 1.197 . The similar
damping coefficients make the demand spectra for La = 0 m and La = 6 m almost identical. The
figure shows that prior to riverbed scouring, the ductility demand caused by MCER is µ dem =1.10 ,
which is 42% smaller than the ductility capacity of µcap =1.89 . The wall pier is design withsatisfactory seismic performance. When the scour depth reaches La = 6 m , the ductility demand
increases to µ dem =1.58 , which is only 9.7% smaller than the ductility capacity of µcap =1.75 .
Riverbed scouring considerably deteriorates the seismic performance of the wall pier bent.

Fig. 3. Seismic performance assessment of the wall pier bent
5. Acknowledgements
This research was sponsored by the Ministry of Science and Technology of Taiwan, R.O.C., underGrant No. 110-2625-M-005-003. The financial support is gratefully acknowledged.
6. References
American Concrete Institute (ACI) 2022. Building Code Requirements for Structural Concrete andCommentary, ACI318, Farmington Hills, Michigan, U.S.A.
American Society of Civil Engineers (ASCE) 2017. Minimum Design Loads and Associated Criteria
for Buildings and Other Structures, ASCE7-16, Reston, Virginia, U.S.A.
Applied Technology Council (ATC) 1996. Seismic Evaluation and Retrofit of Concrete Buildings.
ATC-40, Redwood City, California, U.S.A.
Ministry of Transportation and Communications (MOTC) 2020. Seismic Performance Assessment
and Retrofit Design Specifications for Highway Bridges, Taipei, Taiwan, R.O.C.
National Earthquake Hazards Reduction Program (NEHRP) 2005. Improvement of Nonlinear Static
Seismic Analysis Procedures, FEMA 440, Washington, D.C., U.S.A.

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Experimental Techniques

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Flexural Behavior of 3D-Printed Beam Members
In-Hwan Yang 1*, Quang-The Bui 2, Jihun Park 2, Jungwoo Lee 3, and Changbin Joh 3
1: Kunsan National University, Kunsan, South Korea; email: ihyang@kunsan.ac.kr: Kunsan National University, Kunsan, South Korea
3: Korea Institute of Civil Engineering and Building Technology, Goyang, South Korea
2*: corresponding author

Keyword: 3D printed mortar; flexural behavior; interlayer reinforcement; rebar
Abstract: 3D printing, which is a promising technology in the construction field, has been
rapidly developing in recent years. Because of the layer–by–layer production process, most of
3D–printed structures are applied to the structure under compression or covers like columns andwalls. Therefore, this study was conducted to investigate the flexural behavior of 3D–printed
beam members. Additionally, some methods to improve the flexural behavior of the 3D–printed
beams were performed. To reach these aims, four different beam members were printed andexamined under 4–point loading test. The effects of interlayer reinforcement, penetrating into
mortar layers with hole of beams, and longitudinal rebars on the bending behavior of 3D–beammembers were examined. The failure pattern of the 3D–printed beams was observed, and theperformance of interlayer reinforcement was evaluated. The 3D–printed beams without
longitudinal rebar were abruptly collapsed at the interlayers of the beams. The flexural capacityand post-cracking behavior of the 3D–printed beams reinforced with rebar were significantly
improved. In addition, the improvement of flexural behavior by adding interlayer reinforcement
to the 3D–printed beam was presented. The addition of the interlayer reinforcement enhanced theflexural capacity by approximately 25% and improved the ductility of the beams.
1. Introduction
Nowadays, the application of 3D printing technology has been paid more attention to theconstruction field because of its reducing construction time, construction without formwork,
design freedom, etc. Due to the layer-by-layer constructing process, bonding strength atinterlayer of the 3D printed structure is considered as the strength of the structure. Additionally,
the 3D–printed structure has the anisotropy properties (Joh et al., 2020; Wolfs et al., 2019)
thereby most 3D–printed structures were applied as wall, column, or structure subjected tocompression (Kreiger et al., 2019). To improve strength of 3D–printed structure in tension,
reinforcement for 3D–printed structure is necessary. According to the previous studies, theflexural strength and post–cracking behavior of the 3D printed mortar were improved
significantly by using the reinforcement for the 3D printed mortar. Although several studies wereconducted to investigate the properties of 3D–printed mortar (Ma et al., 2019; Park et al., 2021),
the evaluation of 3D–printed mortar in large scale is still limited. Therefore, this study was
performed to examine the flexural behavior of the 3D–printed beam subjected to flexure as wellas the feasibility of reinforcement for the 3D-printed structure.

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2. Mixtures
The mixture used for the 3D–printing process is given in Table 1. In this study, Ordinary
Portland cement (OPC) with a specificity of 3.14 g/cm3 was used as a cementitious material. To
improve the extrusion and bonding properties, silica fume (SF) with a SiO2 content of 91.3% wasused as a supplementary cementitious material. Class C Fly ash (FA) used in the mixture had a
specific gravity of 2.26 g/cm3 and a loss on ignition of 2.8%. Sand with a size range from 0.160.2 mm was selected to ensure good extrusion during the printing process. To increase theconcrete filament extrusion under the low water-binder ratio of 0.29, a high-performance waterreducing agent (HWRA) was added to the mixture. Finally, the mixture included a viscosity
agent to prevent segregation of the materials and enhance the viscosity of the mixture.
W/B
(%)
0.29

Table 1. Mixture of the 3D printing mortar
Unit weight (kg/m3)
Water OPC SF FA Sand HWRA Viscosity Agent
240

576

79

172

1154

8.27

1.65

Note. OPC: ordinary Portland cement, SF: silica fume, FA: fine aggregate, and HWRA: highperformance water-reducing agent.
3. Experimental Program
The beam dimensions are shown in Figure 1. Four beams were printed by layer with a thick of
10 mm. To prevent the buckling failure during the printing process, the breaking time of 30
minutes was set per each 30 printed layers. Table 2 lists the parameter of the beam member.

Fig. 1. Beam dimensions
Table 2. Parameter of beam members
Interlayer
Filling concrete
Beam namereinforcement
L1-F0-R0
L1-F1-R0
L0-F1-R1
L1-F1-R1

Φ3mm(SL=40mm)
Φ3mm(SL=40mm)–
Φ3mm(SL=40mm)

–
–
Yes
Yes

Rebar
–
–
2Φ13mm
2Φ13mmTo improve the tensile strength of the 3D–printed beams, steel interlayer reinforcements with adiameter of 3 mm and a spliced length of 40 mm were penetrated into the printed layers of the
L1-F0-R0, L1-F1-R0, and L1-F1-R1 beams. In addition, the L0-F1-R1 and L1-F1-R1 beams
were reinforced by 2 steel rebars with a diameter of 13 mm and the filling concrete. To
investigate the flexural behavior of the printed beams, a four-point bending test was performed.
The printed beam was put on a pair steel roller supports with a span length of 900 mm. A
displacement control speed of 1.5 mm/min was applied in this bending test. The actual bending260

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beam test is exhibited in Figure 2. One LVDT was set under the beam to measure thedisplacement during the test.

Fig. 2. 4–point bending test
4. Flexural Behavior
The failure pattern of the 3D–printed beam with and without rebar and filling concrete arepresented in Figure 3 and 4. It shows that the 3D–printed beams without rebar and fillingconcrete were collapsed by the abrupt separation of middle section. Meanwhile, by reinforcing
rebar and filling concrete, this failure was prevented definitely in the L0-F1-R1 and L1-F1-R1
beams. These beams were failed by a shear crack occurring from the support to the loading point.

Fig. 3. Failure pattern of 3D printed beam
without rebar and filling concreteFig. 4. Failure pattern of 3D printed beam withrebar and filling concrete

Figure 5 illustrates the load-deflection curves of four 3D–printed beams.
160

L1-F0-R0 beam
L1-F1-R0 beam
L0-F1-R1 beam
L1-F1-R1 beam

140

Load (KN)

120
100
80
60
40
20
0

0

2

4

6

8

10

Deflection (mm)1214Fig. 5. Load-deflection curves

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The test results show that the L1-F0-R0 beam, including only interlayer reinforcement presented
the lowest load capacity of 21.66 kN. By including interlayer reinforcement and filling concrete,load capacity of the L1-F1-R0 was enhanced by 46.2% with a value of 31.67 kN. The superior
improvement in flexural behavior were exhibited in the L0-F1-R1 and L1-F1-R1 beams
including rebar and filling concrete. For the L0-F1-R1 beam, the maximum load was 114.37 kN
and the ultimate deflection of this beam was 14.36 mm. It implied that the load capacity of the
3D–printed beam was improved by approximately 5 times by using rebar and filling concrete in
the hole of the beam. The largest load capacity of 142.22 kN was obtained in the L1-F1-R1beam,
which was enhanced by interlayer reinforcements, rebar, and filling concrete. It indicates thatwith the addition of interlayer reinforcement, the load capacity of the 3D–printed beam was
improved by approximately 25% when compared the L0-F1-R1 and L1-F1-R1 beams. In
addition, the deflection of the beams including rebars and filling concrete were higher than thatof the beams without rebar.
5. Conclusions
The flexural behavior of the 3D–printed beam was evaluated in this study, and some conclusions
could be drawn as follows:• The failure pattern of 3D–printed beam implies that the bonding strength between theinterlayers played importance roles in flexure. Therefore, the reinforcement is required
for the 3D–printed member.
• The addition of the interlayer reinforcements enhanced the load capacity of the 3D–
printed beam by approximately 25%.
• The flexural behavior of the 3D–printed concrete was improved significantly by using
rebar and filling concrete.
• By combining interlayer reinforcement, rebar and filling concrete, the L1-F1-R1 beam
exhibited the superior performance of the flexure when compared to the others.
6. References
Joh, C., Lee, J., Bui, T. Q., Park, J., & Yang, I.-H. (2020). Buildability and Mechanical
Properties of 3D Printed Concrete. In Materials (Vol. 13, Issue 21).
https://doi.org/10.3390/ma13214919.
Kreiger, E. L., Kreiger, M. A., & Case, M. P. (2019). Development of the construction processes
for reinforced additively constructed concrete. Additive Manufacturing, 28, 39–49.
https://doi.org/https://doi.org/10.1016/j.addma.2019.02.015.
Ma, G., Li, Z., Wang, L., & Bai, G. (2019). Micro-cable reinforced geopolymer composite for
extrusion-based 3D printing. Materials Letters, 235, 144–147.
https://doi.org/10.1016/j.matlet.2018.09.159.
Park, J., Bui, Q.-T., Lee, J., Joh, C., & Yang, I.-H. (2021). Interlayer Strength of 3D-Printed
Mortar Reinforced by Postinstalled Reinforcement. Materials, 14(21), 6630.
https://doi.org/10.3390/ma14216630.

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Wolfs, R. J. M., Bos, F. P., & Salet, T. A. M. (2019). Hardened properties of 3D printed
concrete: The influence of process parameters on interlayer adhesion. Cement and Concrete
Research, 119, 132–140. https://doi.org/10.1016/j.cemconres.2019.02.017.

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ELSS Joint: Development of Semi-Rigid Joint for Joining Precast Prestressed
Concrete Slabs in Japan
Mohammad Emran Nasery1*, Tatsuhiko Mimoto2, and Nobuaki Sakurai3
1

Kyokuto Kowa Corporation, Hiroshima, Japan; email: Nasery@kkn.co.jp
Kyokuto Kowa Corporation, Hiroshima, Japan; email: Mimoto@kkn.co.jp
3
Nippon Steel Engineering, Tokyo, Japan; email sakurai.nobuaki.xd8@eng.nipponsteel.com
2*

Corresponding author

Keywords: Precast concrete slabs; Joint; Loop joint; ELSS joint; Productivity; Slab replacement
Abstract: Approximately 40% of the expressways in Japan have been used for 30 years or more.
These expressways have deteriorated due to aging, excessive traffic load, and environmental
factors. Large-scale renewal and repair projects, which include bridge slab replacement, havebeen launched all over Japan. Because the expressways are in operation, slab replacement workneeds to be done as quickly as possible. However, the commonly used loop joints for joining
precast prestressed concrete slabs do not allow the construction time to be shortened due to theirstructural complexity. This study evaluates the structural durability of a newly developed semirigid joint, called the Economy, Labor Saving, Speedy (ELSS) joint, and its advantages over thecommonly used loop joints in terms of construction time and efficiency (labor man-hours). To
confirm the structural durability of the ELSS joint, the results of a wheel load running test arereported. Furthermore, a loop joint and the ELSS joint are compared in terms of material usage
and construction time using an example. The results show that for joining precast prestressed
concrete slabs, the ELSS joint provides sufficient structural durability, shortens construction time,
reduces labor man-hours and material usage, and simplifies future slab replacement work
compared to loop joints.
1. Introduction
The majority of the roads and bridges in Japan were constructed during a period of higheconomic growth after World War II. Approximately 40% of Japanese expressways have beenused for 30 years or more. These expressways have deteriorated due to aging, fatigue loading
caused by heavy traffic, and environmental factors such as damage from salt (used as an antifreezing agent) (Kita et al., 2020). To address this challenge, Nippon Expressway Company(NEXCO), which is the main operator of expressways in Japan, has launched large-scale renewal
and repair projects. As part of these projects, the renewal, repair, and reinforcement of highway
bridge slabs are being carried out. Slab replacement work has thus increased all over Japan. The
replacement work, which is expected to cover approximately 230 km, is mostly for reinforced
concrete (RC) slabs that have deteriorated due to fatigue.
The completion of these large-scale renewal projects faces some challenges. NEXCO
aims to complete these projects with minimal traffic closure, efficient replacement work,
improved durability of the slabs (slabs should have a life span of 100 years), and simplified
future maintenance work. Thus, precast prestressed concrete slabs are the best option in terms of264

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construction time and durability. Precast slabs are commonly joined using RC joints called loop
joints, as shown in Fig. 1. However, loop joints have several limitations, such as the required
formwork, complex rebar arrangement, inability to be used for thin slabs, required concrete
casting and curing, lengthy construction time, difficult future replacement (loop joint rebar mustbe cut), and need for several skilled workers.
To address these challenges, Nippon Steel Engineering and Kyokuto Kowa Corporation
have collaboratively developed a new semi-rigid joint, called the Economy, Labor Saving,
Speedy (ELSS) joint, that has sufficient durability while reducing the construction time andsimplifying the construction method compared to those for loop joints. The ELSS joint is shownin Fig. 2.
The ELSS joint is the result of several experiments conducted to determine the best joint
filler materials, required thickness and shape of the joint, and structural performance anddurability under static and dynamic loading. From the results of several experiments andanalyses, it was concluded that if (i) slabs are additionally reinforced with rebar in areas adjacentto the slab-joint interface, (ii) the joint filling material has a Young’s modulus of 1/20 to 1/3 ofthat of the slab concrete, (iii) the joint has a width of at most 40 mm, (iv) a joint filler material
with a bond tensile strength equivalent to or greater than that of polymer cement mortar or epoxy
resin mortar is used, and (v) the bond tensile strength is equal to or greater than the tensile
strength of the concrete in the slab, then the bond shear strength at the slab-joint interface will be
sufficient to prevent slab-joint peeling or detachment and the vertical displacement will not cause
any degradation of the water resistance and asphalt durability (Kita et al., 2017; Sakurai et al.,2018; Kita et al., 2020).
Precast slab

Precast slab

Joint filler

Rebar

ELSS joint

Loop joint

Fig. 2. Diagram of ELSS joint

Fig. 1. Diagram of conventional loop joint

2. ELSS Joint Characteristics and Advantages
The ELSS joint has a simpler construction method than that for loop joints as there is no rebar in
the joint, which greatly reduces the joint casting time and future replacement work. Moreover, as
starter bars coming out of the precast slabs are not necessary for the ELSS joint, the formwork
and rebar arrangement for the precast slabs in the factory can be simplified, improving
manufacturing productivity. The maximum transportable panel (slab) size in Japan is 2.5 m. The
slabs designed for loop joints have a maximum transportable width of 2.1 m and those for the
ELSS joint have a maximum transportable width of 2.4 m.
350

Allocated slab width
2100 mmTransportable width
2383 mm

30

350
(a)

Allocated slab width
2430 mm30

Transportable width2400 mm(b)

Fig. 3. Cross-sectional view of (a) loop joint and (b) ELSS joint
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This longer maximum width can reduce the number of slabs that need to be designed andmanufactured for the ELSS joint by about 15% on average, which will shorten the slab
manufacturing time in the factory and replacement activity on site. Figure 3 shows crosssectional views of a loop joint and the ELSS joint.
3. Wheel Load Running Test for Fatigue Durability of Slab-ELSS Joint Connection
According to the Specifications for Highway Bridges of the Japan Road Association, the RC
slabs need to be examined and verified for fatigue durability through step loading. To achievethis, a running wheel load of 250 kN was run over the slab back and forth 100,000 times and thedeflection and cracks at the bottom face of the slab were evaluated (Goto et al., 2020). The
durability was then judged by checking whether there was any water leakage on the bottom faceof the slab when water was pooled on the top of the test specimen. To verify the fatigue
durability of the ELSS joint, we conducted a wheel load running test (NEXCO test method 4422019) on two slabs connected by an ELSS joint (joint width: 25 mm). The two specimens wereconstructed as per NEXCO test method 442-2019. The ELSS joint materials used for these
specimens were polymer cement mortar and epoxy resin mortar, respectively. The results showthat there was no water leakage in the test specimens where negative bending was not present(the top face of the slab-joint connection was in compression). This means that the ELSS joint
can be used in NEXCO projects for slab-joint connections for which the top face is incompression without any safety concerns.
4. Comparison of Loop Joints and ELSS Joint in Actual Bridge Construction Project
To compare the differences in using the two types of joints, a three-span continuous steel noncomposite slab girder bridge with an overall length of 123.1 m (each span had a length of 40.7 m
and a total width of 10.66 m) was selected. The RC slabs were to be supposedly replaced withprestressed concrete slabs. Table 1 presents the material usage for the loop joints and the ELSS
joint for this project. As shown, there is a drastic difference in material usage; the ELSS will
reduce the project cost and shorten the construction time. The number of precast slabs required
for loop joints and the ELSSS joint is 61 and 50, respectively. This means that a shorterproduction time is required for precast slabs when the ELSS joint is used. Additionally, thereduction in material, rebar, and formwork usage for the ELSS joint is substantial.
Table 1. Comparison of using loop joint and ELSS joint
Total number of precast slabs
Joint
Materials
Rebar
Formwork

Loop joint
61
49.4 m3
8.851 tons
201 m2ELSS joint
50
4.6 m30.000 tons
36 m2

Difference↓11↓44.8 m3
↓8.851 tons
↓165 m2

Furthermore, Table 2 and Fig. 4 demonstrate that using the ELSS joint can reduce the overall
construction time by 11% and the labor man-hours by 14%, particularly in slab manufacturing
and joint construction.

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Table 2. Construction time comparison of loop joints (blue) and ELSS joint (orange)
Activity

Traffic operation switching
Removal of existing floor slab /temporary precast floor slab
Precast floor slab joint
Bridge construction work
Traffic operation switching

1st Month
10 20 30

2nd Month
10 20 30

3rd Month
10 20 30

4th Month
10 20 30

11% reduction

Slab manufacturing
Old slab removal, new slab installation
Joint construction

14% reduction

Fig. 4. Productivity (man-hours) comparison of loop joints and ELSS joint
5. Conclusion
This study evaluated the characteristics of the ELSS joint and its advantages over theconventionally used loop joints for joining prestressed concrete bridge slabs in Japan. The results
show that using the ELSS joint reduces the construction time, material usage, and labor manhours and simplifies future replacement work compared to using loop joints, while providing therequired structural performance and durability. In future research, we will expand the usage of
the ELSS joint that is subject to different types of loading. Particularly, we are trying to findways to achieve a desirable result when using the ELSS joint in slab-joint connections subjected
to a negative bending moment (the top face of the slab-joint connection is in tension).
6. References
Goto, S, Nagatani, T, Homma, A & Hirano, K (2020) ‘Proposal of fatigue durability evaluation
method for PC slabs’ Journal of Structural Engineering, Japan Society of Civil Engineers Vol
66A, pp. 762-773,
Kita, S, Nakayama, H, Sakurai, N, Maekawa, K, Maruno, M, Satake, S, Oigawa, Y & Taniguchi,
S. (2017) ‘A study of strength evaluation of a joint structure using a beam model in semicontinuous precast floor slabs’, Journal of Structural Engineering, Japan Society of Civil
Engineers, Vol 63A, pp.49-56.
Kita, S, Sakurai, N, & Maekawa, K (2020) ‘Durability Evaluation with Wheel Running Tests
and Computational Simulations on Precast PC Slab with Half Rigid Joint Under the Negative
Bending Moment’. Journal of Japan Society of Civil Engineers, Ser. E2 (Materials and Concrete
Structures), 76(4), pp.403-420.
NEXCO Test method, Volume 4, (2019) Structural test method, 442-2019, P.91-94
Sakurai, N, Maekawa, K & Matsui, S (2018) ‘Semi-rigid Connections for Precast Concrete Slabs,
Bridges and Foundations’, Vol.52, No.9, pp.29-36,

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Horizontal–Vertical Ratio for Concrete Pumping Pipe
Ki-Yeol Kim1, Young-Jin Kim2, and Myoung-Sung Choi3*1: Dankook University, Youngin, South Korea; email: kiyeol520@naver.com: Korea Concrete Institute, Seoul, South Korea; email: kimcrete@naver.com
3: Dankook University, Youngin, South Korea; email: choims@dankook.ac.kr
2*: corresponding author

Keywords: Concrete pumping; Pipe; Rheology; Horizon-vertical ratio; Computational fluid
dynamics
Abstract: For the construction of high-rise building, a horizontal pipe pumping test is usuallyconducted and using obtained test results, vertical actual constructing condition could be
evaluated. In this study, horizontal and vertical pumping experiments are performed undervarious mixing and construction conditions using rheology-based quantitative performance
prediction technology to derive a horizontal–vertical ratio of concrete pumping pipe for
evaluating concrete pumping performance. Computational fluid dynamics is used to analyze thecorrelation between experimental results and additional vertical lengths. Analysis results confirm
that compressive strength and plastic viscosity correlate with vertical and horizontal correlation
coefficients; the pipe length, flow rate, and yield stress have no significant effect on thecorrelation coefficient. Based on analysis result considering various factors, multiple regression
analysis is performed to propose horizontal–vertical ratio relationship for the concrete pumpingpipe.
1. Introduction
Concrete pumping is used to transport concrete to a target location using a pump and piping. The
concrete transport via pumping is highly efficient in terms of the amount of poured concrete,pouring time, and cost compared to methods that use buckets, conveyors, and chutes. Further,
concrete pumping can ensure even pours at sites where entry of large equipment is restricted ordifficult. Thus, high-rise buildings are constructed using concrete pumping at most construction
sites.
To this end, a quantitative prediction method for pumping performance was developed
recently based on rheology; this method has applied to actual construction sites. Other studies
have investigated the concrete pumping mechanism and influencing factors, and various basic
technologies required for pumping design have been developed based on such studies.
Thus, in this study, pumping tests for horizontal and vertical pipes were conducted undervarious mixing and construction conditions. We used the rheology-based quantitative pumping
performance prediction technology to overcome the limitations of the existing horizontal–
vertical ratios. The test results and computational fluid dynamics (CFD) analysis results wereanalyzed to propose a new horizontal–vertical ratio of the concrete pumping pipe.

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2. Pumping Tests
2.1 Material properties
Concrete mixtures used in the vertical and horizontal pipe pumping tests include the generalstrength concrete of 24, 27, and 30 MPa and high-strength concrete of 40, 60, and 80 MPa; thesemixtures are commonly used in the field. Table 1 lists the materials and the mix proportions foreach strength. Ordinary Portland cement with a specific gravity of 3.15 was used. For fineaggregates, natural river sand with a particle size of 0.08 ~ 5 mm was mixed with crushed stone
at a 7:3 ratio. The fine aggregate had a specific gravity of 2.59, fineness modulus of 2.81, and amoisture content of 2.43%. A coarse aggregate made of limestone with a maximum size of 20
mm was used, and it had a specific gravity of 2.61, fineness modulus of 5.65, and moisture
content of 0.8%. The mixing water was adjusted by considering the moisture content of the fineand coarse aggregates. A PC-based high-range water-reducing admixture (HRWRA) was used,
and the amount used was shown as a percentage (%) compared to the weight of the binder listed
in Table 1.
Table 1. Mix properties used for horizontal and vertical pipe pumping test
Design strength
Normal strength (MPa)High strength (MPa)Material
24
27
30
40
60
80
Ordinary portland cement (kg/m3)
245
263
287
282
297
389
Water-binder ratio
0.48
0.45
0.40
0.35
0.32
0.25
Fly ash (kg/m3)
35
38
41
47
54
33
3
Blast furnace slag (kg/m )
70
75
82
141
162
195
3
Slica fume (kg/m )33
Gravel (kg/m3)
921
900
910
961
958
891
Sand (kg/m3)
905
884
828
751
696
696
High-range water-reducing admixture (%)
0.35
0.31
0.28
0.18
0.24
0.25
2.2 Experimental pipe
Pumping tests for horizontal and vertical pipes are conducted to obtain the vertical and horizontal
correlation coefficients for concrete pumping, as shown in Figs. 1. The horizontal pipe pumping
test was conducted using pipe lengths of 100, 200, 500, and 1000 m, as shown in Fig. 1(a). In
this test, pressure gauges were installed 10 m from the pump outlet and at 3 m before the outlet
for each target pumping length to measure the pressure. For the curved pipe, a 90◦ curved pipewith a radius of 1.0 m was used. The number of installed curved pipes was minimized to reduce
their influence because there has been considerable controversy over the influence of curved
pipes on the pressure drop. Fig. 1(a) shows the pressure measurement locations for the entire
length of the horizontal pipe pumping test.
Fig. 1(b) shows the vertical test was conducted using vertical pipe lengths of 100 and 200 m.
In this test, the first pressure gauge was installed 20 m from the pump inlet in the horizontal
section, and the last pressure gauge was installed 3 m before each outlet in the vertical section,
similar to that in the horizontal pipe pumping test. For the curved pipe, a 90◦ curved pipe with aradius of 1 m was used. One curved pipe was installed at the location of the first pressure sensor,
and the other at the outlet. All pipes had a diameter of 127 mm and a thickness of 7.7 mm.

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(a)

(b)

Fig. 1. Concrete pumping test: (a) Schematic ground plan of the horizontal pumping circuits andthe location of the pressure gauges; (b) Schematic ground plan of the vertical pumping circuits
and the location of the pressure gauges
In the tests, the flow rate was calculated considering the filling coefficient by adjusting the stroketime of the pump under various mixing conditions; the pressure was measured using the pressuregauges installed in the pipes.
3. Result of Analysis
3.1 Prediction of vertical pumping performance using the CFD technique
The experimental and analysis results for vertical heights of 100 and 200 m were compared toverify the analytical technique used;the results are shown in Fig. 3. The results show that the
CFD technique is effective under the experimental conditions of this study.Therefore, it is worth
to apply for the additional lengths of 500 m and 1000 m. The experimentally determined
thickness of the lubrication layer and the rheological constants of the concrete and lubrication
layer were used as input data, and the pressure required for the target flow rate at each verticallength was determined and utilized for analyzing the horizontal–vertical ratio.
140

70

Experimental Data
CFD Analysis Result

50
40
30
20

100
80
60
40
20

10
0

Experimental data
CFD analysis result

120

Pressure (bar)Pressure (bar)

60

24

27

30

40

60

Compressive Strength (MPa)

0

80

24

27

30

40

60

80

Compressive strength (MPa)(a)(b)
Fig. 2. Comparison between experimental data and CFD analysis: (a) 100 m case; (b) 200 m case

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3.2 Horizontal–vertical ratio
The pumping tests for the horizontal and vertical pipes and CFD analyses were conducted undervarious compressive strengths, pipe lengths, and flow rate conditions; the horizontal–verticalratios were derived. The analysis results indicated that the compressive strength and plastic
viscosity are closely related to the horizontal– vertical ratio, and the pipe length, flow rate, andyield stress have no significant influence on the horizontal–vertical ratio. Based on the analysisof the experimental results derived in this study, multiple regression analysis was conducted foreach variable set; a relationship for the horizontal–vertical ratio was proposed as=
Y 4.8 − 0.73µ L(1)where Y denotes the horizontal–vertical ratio, and µ L denotes the plastic viscosity of the
lubrication layer. That is, the horizontal–vertical ratio is directly affected by the plastic viscosityof the lubrication layer in the same manner as the concrete pumping performance.
5500 m
R2=0.64

4

Horizontal-vertical ratio

Horizontal-vertical ratio

5

3
2
1
0

0

20

40

60

80

4
3
2
1
0

100

Compressive strength (MPa)

Flow rate : 40 (m3/h)
R2=0.76

0

20

40

60

80

100

Compressive strength (MPa)(a)(b)
Fig. 3. Variation of horizontal-vertical ratio: (a) according to pipe length; (b) according to flowrate
4. Conclusion
In this study, horizontal and vertical pipe pumping tests and computational fluid dynamics (CFD)
analysis were conducted under various mixing and construction conditions using rheology-based
quantitative pumping performance prediction technology to derive the vertical-horizontal
correlation used for concrete pumping performance evaluation.
5. Acknowledgements
This work is supported by the Korea Agency for Infrastructure Technology Advancemen(KAIA)
grant funded by the Ministry of Land,Infrastructure and Transport (Grant RS-2021-KA163381).
6. References
Choi, M.S. 2013. Prediction of concrete pumping performance base on the evaluation of
lubrication layer properties, Korea Adv. Inst. Sci. Technol.

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Kwon, S.H., Park, C.K., Jeong, J.H., Jo, S.D., and Lee, S.H. 2013. Prediction of concrete
pumping: part i-development of new tribometer for analysis of lubricating layer, Journal of ACI
Materials, 110, 647–655

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Over Torque Test Method to Assess the Concrete Strength
Kenji Tada*1, Tatsuki Hatakeyama 2, and Tetsuya Ohmura 3
1: JUST Co. Ltd., Kanagawa, Japan; tada@just-ltd.co.jp: Tokyo City University, Tokyo, Japan; g2281646@tcu.ac.jp
3: Tokyo City University, Tokyo, Japan; tohmura@tcu.ac.jp
2*: corresponding author

Keywords: torque; concrete strength; concrete screw; drilling; screw-in
Abstract: Recently, the demand for seismic assessment, structural survey, and quality control has
further increased. The top priority is the concrete strength for assessment of the concrete structure.
Generally, some concrete samples are taken to test the concrete strength, however, that needs muchcost, labor, work period, and the heavy-duty equipment of concrete core drill. In addition, theconcrete strength of the concrete samples must be tested in a laboratory after taking the concrete
samples. On the other hand, the rebound hammer method, so-called the ‘non-destructive test,’
would sometimes be used, though the accuracy of the concrete strength was suspicious. And apublic assessment with the non-destructive test is not accepted in Japan because of the accuracyof the test. In this study, we developed the over-torque time integration test method using a small
drill. The method is independent of the worn condition of the drill bit and the drilling pressure. In
the method, at first, a small guide hole is drilled. Second, a small concrete screw is slowly tightened
to the head with a manual screwdriver. Then, the small concrete screw is driven while measuring
the torque. The torque measured at this time is called over-torque. Finally, the concrete strength isassessed compared with the concrete strength and the over-torque-time property relationship. A
lot of the over-torque test methods were performed while measuring the torque, and the test results
were examined. The torque was steeply increased at the start screw-in. Next, the torque is kept
while driving the drill to some extent. Finally, the torque steeply dropped down. Just a drilled small
hole was left. The relationship between the concrete strength and over-torque-time property was
shown and we confirmed that the over-torque test method could be highly valuable to assess theconcrete strength.
1. Introduction
Large earthquakes, for example, the Kobe earthquake in 1995, Niigata Prefecture Chuetsu
Earthquake in 2004, the Great East Japan Earthquake in 2011, and the Kumamoto earthquake in2016, have occurred, and the subsequent colossal earthquake could hit Tokyo in the near future.
Recently, the demand for seismic assessment, structural survey, and quality control has increasedin that background. The well-known key thing is the concrete strength for an assessment of aconcrete structure.
Generally, some concrete samples are taken to test the concrete strength, however, thatneeds much cost, labor, work period, and the heavy-duty equipment of concrete core drill. In

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addition, the concrete strength of the concrete samples must be tested in a laboratory after taking
the concrete samples.
On the other hand, the rebound hammer method, so-called the ‘non-destructive test,’ would
sometimes be used, though the accuracy of the concrete strength was suspicious. And a public
assessment with the non-destructive test is not accepted in Japan because of the accuracy of thetest.
In a past study, a small drilling test for concrete strength was reported focusing on the
average consumed power while drilling. However, the test result might depend on the worncondition of the drill edge and the drilling pressure.
In this study, we developed the over-torque time integration test method using a small drill.
The method is independent of the worn condition of the drill bit and the drilling pressure. In themethod, at first, a small guide hole is drilled. Second, a small concrete screw is slowly tightened
to the head with a manual screwdriver. Then, the small concrete screw is tightened whilemeasuring the torque. The torque measured at this time is called over-torque. Finally, the concrete
strength is assessed compared with the concrete strength and the over-torque-time integral value.
2. Test2.1. Test specimen
Figure 1 shows the test specimen. Specimens 100mm in diameter and 200mm in height wereprepared at first and cut at the center of the height next. Specimens were tested on the mid-cut faceof the lower specimen. Three same specimens were made for each of the 16 different concrete
strengths., and a total of forty-eight specimens were tested.
Their concrete strength f ’c was in the range of 4.84 to 94.2 N/mm2 for specimens. The maximum
aggregate size was 20mm.
2.2. Test setup and measurement
Photo 1 shows the test setup. The telemeter to measure the torque was first jointed between thescrewdriver bit and the driver chuck. Next, a computer was connected to the telemeter via the datalogger.

PC

Cut to two pieces

200mm

100mm
Upper

Test surface

Used for testing
Lower

Drill
Telemeter
LoggerPhoto 1 Test setup

Fig. 1 Specimen274

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20mm

15mm

Measurement point
TV15

Fig. 2 Measurement position

30mm

Test surface

TV20

TV30

Photo 2 The test screwFirstly, the guide hole was drilled without measurement. Second, the test screw is slowly tightened
to the head with a manual screwdriver. Then, the test screw is tightened while measuring the torque.
The test was stopped when the torque was reduced to zero. Photo 2 shows the test screw. For thetest screw, a hexagonal bit was braze welded to the concrete screw head with a length of 35 mm
under the neck and a shaft diameter of 7 mm. Forty-eight test screws were prepared for each of thethree types of test screws, with the length of the screw blade part from the tip 15, 20, and 30mm.
Figure 2 shows the measurement position. The vertex of an equilateral triangle with the center ofgravity at the center of the test surface was used as the measurement point. A new test screw wasused for each measurement point, and the measured value of one specimen was the average valueof the measured values measured at three measurement points.
3. Test Results
3.1 Maximum torque and concrete strength
The torque was steeply increased for all specimens when the test screw started to be screwed in.
Finally, the torque drastically dropped to zero after repeatedly increasing and decreasing for awhile. Table 1 shows the test results. N/A was indicated to be impossible to measure the torquebecause of the test screw defect. TV 15, 20, and 30 indicate the length of the screw blade part.
Figure 3(a) shows the maximum torque and concrete strength. Three regression lines by the leastsquares method indicated the lengths of the screw blades 15, 20, and 30 mm. Maximum torquetends to increase as concrete strength increases.

Table 1. Test results
No.
①
6 ②
③
①
7 ②
③
①
8 ②
③
①
9 ②
③
①
10 ②
③

f 'c(N/mm²)

4.84
10.2
15.2
19.8
22.5

Maximum torque (N・m）

TV15

TV20

TV30

0.232
0.206
0.371
0.220
0.489
0.577
0.766
1.071
1.166
1.217
1.316
0.681
1.243
1.363
0.611

0.146
0.412
0.187
0.588
0.976
0.605
1.405
1.268
0.280
0.983
1.101
1.845
1.464
2.074
1.247

1.468
1.083
1.404
1.321
1.485
1.365
2.069
1.743
1.547
2.168
1.155
2.126
2.263
1.444
1.612

No.
①
11 ②
③
①
12 ②
③
①
13 ②
③
①
14 ②
③
①
15 ②
③

f 'c(N/mm²)

24.1
26.7
29.2
39.2
61.2

Maximum torque (N・m）

TV15

TV20

TV30

0.860
1.077
1.544
2.104
0.939
0.883
1.237
0.761
0.642
1.930
1.943
－
1.869
2.385
2.552

1.588
2.074
1.397
2.114
1.613
1.309
1.324
1.331
1.005
2.351
1.286
1.328
2.360
1.988
2.751

2.100
1.668
1.600
1.775
2.229
1.560
2.370
1.789
1.748
－
1.726
1.520
2.867
2.843
2.854

TV15, 20, and 30 indicate the length of the screw blade part.

275

No.
①
16 ②
③
①
17 ②
③
①
18 ②
③
①
19 ②
③
①
20 ②
③
①
21 ②
③

f 'c(N/mm²)

72.4
81.1
82.7
89.9
90.8
94.2

Maximum torque (N・m）

TV15

TV20

TV30

2.378
1.591
1.875
2.295
2.763
2.661
2.791
2.895
－
1.852
1.877
2.339
1.597
2.186
2.078
1.355
－
－

2.771
1.905
2.577
2.637
2.257
2.728
2.823
2.807
2.831
2.200
2.122
1.687
1.765
1.841
1.943
1.843
1.751
－

2.820
2.881
2.823
－
3.087
2.285
2.934
2.958
－
－
2.288
－
－
2.462
－
2.813
－
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Bridge Engineering Institute Conference 2023 (BEI-2023)
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3.2 Torque time history
Figure 3(b) shows an example of the torque time history. It approximately tends to be increased
following an increase in the concrete strength.
The torque was steeply increased at first, and the torque was kept for a couple of seconds next andfinally dropped down to zero.
The torque time history property was influenced by the concrete strength and the coarse aggregate
location.
3.3 Torque time integral value and concrete strength
Figure 3(c) shows the integral torque time value and concrete strength. The dot lines respectively
indicated the regression lines.
The integral torque time value was increased following the increase in the concrete strength.
The variance is approximately smaller. And the dispersion for the specimen of the length of thescrew blade 20mm was much less than others.
f ′c=100.9N/mm²

TV15
TV20
TV30

Gravel affect

TV30

50.2

TV20

58.8

TV15

(a) Maximum torque andconcrete strength(b) Torque time history(c) Torque time integral value
and concrete strength

Fig. 3 Correlation between each test result and concrete strength

4. Conclusions
We developed the over-torque test method with a small drill. The method was to drill a guide hole
first, slowly tighten the test screw to the head with a manual screwdriver, and slowly screw it inwhile measuring the torque until it reached zero.
The specimen was a concrete cylinder. Since the strength of concrete was considered to be affectedby the settling of aggregates, the test was performed on the cut surface of the lower specimen
obtained by cutting it to half the height.
After reviewing the test results, we came to the following conclusions.
1) The over-torque test method we developed was simple to perform and get the torque data.
2) Torque time history tends to increase as concrete strength increases.
3) The maximum torque was correlated with the concrete strength.
4) The torque time integral value was correlated with the concrete strength.
5) For the standard deviation, the test results for the length of the screw blade 20mm were muchbetter than that of 15 and 30mm.
6) The dispersion was shown as the relationship between the torque and concrete strength.
Therefore, to improve the accuracy, it is necessary to test with a wide variety of concrete andobtain more data.

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Application of Automatic Welding Robotin Integrated Construction of Bridge Pile Foundation
Liangjun Hu 1* , Xiaoli Sun 2, and Jun Yang3
1: Postdoctoral Mobile Station，Tsinghua University，Beijing, China; 2. Guangzhou Instituteof Building Science Company Limited, Guangzhou, China; Guangzhou Municipal
Engineering Testing Co., Ltd., Guangzhou, China; Liangjun_hu@outlook.com
2: Guangzhou Municipal Engineering Testing Co., Ltd., Guangzhou, China;
24620011@qq.com
3: Guangzhou Municipal Engineering Testing Co., Ltd., Guangzhou, China;
yangjun851113@163.com*: corresponding author
Keywords: pile foundation; welding robot; bridge construction
Abstract: Undertaking construction work in the summer can be challenging and draining,
especially for welders who must wear protective gear and operate in high-temperature
conditions. Technological innovations ought to offer workers a more respectable andcomfortable work experience. In line with this, an automatic welding robot has beendeveloped to enhance the efficiency and quality of bridge pile foundation construction.
Prestressed high-strength concrete (PHC) pipe piles are a popular and cost-effective method
used in bridge construction thanks to their high single-pile bearing capacity, strength, andreliability. However, the manual welding of large-diameter (over 600 mm) pipe piles is alow-efficiency and unstable construction process that can compromise the quality of the pilefoundation, thereby affecting the overall quality of the bridge. To address this challenge, thereconstruction and expansion project of the Shenzhen-Shanxi Expressway has developed apile-beam integrated (PBI) machine that utilizes factory prefabrication and mechanizedassembly in-situ to enhance the quality and speed of bridge construction. In addition, tofurther improve construction progress, our research team has developed and employedautomatic welding robots for the connection work of piles, which have been installed in thepiling area of the PBI machine. Our study demonstrates that using automatic welding robots
in large-diameter PHC pipe piles can significantly reduce physical labor and improve theefficiency and quality of bridge pile foundation construction. In addition, the welding robot
has helped achieve an intelligent, rapid, and standardized construction process, thus
contributing to the overall success of the expressway project.
1. Introduction
Prestressed High Strength Concrete Pile (PHC Pipe Pile) has the advantages of high single
pile bearing capacity, high pile body strength, strong ability to penetrate the soil layer, fast
construction speed, low cost per unit bearing capacity[1][2]. The annual usage of prestressed277

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concrete pipe piles exceeded 300 million linear meters in China, which are widely used inhigh-rise buildings, bridges, ports, wharves and other projects.
Connecting prestressed pipe piles using manual welding involves several technical steps,
requires high expertise, and poses health risks due to arc light and radiation exposure.
Additionally, labor-intensive manual welding is time-consuming, impacting the construction
schedule, while the welding industry is facing a shortage of workers and an aging workforce.
To address these challenges, the research team developed a welding robot designed forconnecting prestressed concrete pipe piles by studying the working principles andapplications of existing automatic welding equipment. The paper first presents thefundamental construction of the newly developed pipe pile welding robot. Subsequently, thewelding quality was evaluated through an indoor steel pipe welding test using the robot.
Lastly, the robot was applied to the reconstruction and expansion project of Shenshanxi
Expressway in Guangdong, and its effectiveness in the field was evaluated throughthe welding test of bridge pipe piles.
2. Fundamental Construction of Welding Robot
Fig. 1 illustrates the structure of the prestressed concrete pipe pile welding robot, whichmainly consists of a welding trolley, a control box, welding equipment, steel belt guide rail,
and a mobile storage cart. The welding trolley includes a clamping mechanism, swing device,driving wheel, motor, etc. The control box includes controller, control handle, servo
controller, etc. The welding equipment comprises not only the automatic wire feeder andwelding machine but also other essential welding auxiliary equipment such as the power box,
shielding gas, among others. The walking track adopts a puller-type quick locking
mechanism [3][4], which realizes quick locking and at the same time forces the steel belt toexert radial pressure on the pipe pile. With the help of this force, the track can hold the pipe
pile tightly. The mobile storage cart is mainly used to store the welding trolley, welding
machine, control box and its pipelines, etc., which can realize the transportation andmovement of the whole machine.

(a) welding trolley(b) welding equipment(c) steel belt guide rail

Fig. 1. Fundamental construction of welding robot
3. Indoor Model Test Research
For this experiment, pre-aligned steel pipes and pipe pile end plates with a diameter of 1mwere used as a scaled model of engineering pipe piles. The indoor welding experimentfocused on the three-layer welding process of the welding robot. Before welding, install thesteel belt guide rail and welding trolley to secure the guide rail onto the pipe pile and ensuresmooth operation. The weld depth is 15 mm, with a width ranging from 14.5 mm to 15.5 mm,

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and the maximum axial deviation of the upper and lower pipe piles is 3 mm. The scaled
model tests primarily the three-pass welding process. The three-layer welding processachieves excellent welding effectiveness and high efficiency. Fig. 2 illustrates the welding
process and outcomes of the three-pass layer welding on the machine. The first layer welding
produces a uniform weld seam with minimal curling, meeting the requirements ofback-sealing welding. High current and large swing parameters are applied during the second
filling welding to ensure sufficient support surface for the third covering welding. The thirdlayer welding produces a full, uniform, and continuous weld seam with smooth filling and no
defects such as pores or slag inclusions, satisfying the requirements set forth in the"Technical Standards for Prestressed Concrete Pipe Pile" (JGJ/T 406-2017). The total
welding time for three-layer welding with a single welding robot is 45 minutes, which can befurther improved by controlling two welding robots simultaneously.

(a) welding process(b) first back-sealing welding(c) second filling welding(d) third covering welding

Fig. 2. Indoor model welding test
4. Application of Welding Robot on Bridge Pile
The TJ10 Meilong Super Bridge of Shenshan West Expressway is being reconstructed andexpanded. The stressed concrete high-strength pipe piles have a pile length of 34 m and adiameter of 1000 mm with a wall thickness of 140 mm. The pile arrangement form is
12m+10m+12m, with the upper part being a PRC (prestressed reinforced concrete pile)
pile and the lower part being a PHC pile. Fig. 3(a) shows the pile-beam integrated (PBI)
machine used in the bridge construction, and the pipe pile foundation for the bridge utilizes
the hammering method to sink the pile. The steel belt guide rail and welding trolley are
installed before on-site welding, and the weld width of pipe piles on-site typically ranges
from 12-14 mm. Figs. 3(b) and (c) display the on-site welding process of the welding
robot and the appearance of the weld after welding. The figures illustrates that the automatic
welding seam is even and has a flat appearance. When compared to the timeliness of on-sitemanual welding, the welding process using two workers to weld a single weld usually takes279

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60 minutes, while the welding robot takes at least 50 minutes (including 8-10 minutes ofpreparation time before welding). Additionally, only one welder is required to operate theequipment and monitor the welding process. Although the efficiency of automatic welding
has significant room for improvement compared to manual welding, the welding quality andefficiency have been unanimously recognized by the application units. From the perspectiveof field application, The entire set of welding equipment is heavy, making it challenging tomove on muddy or uneven construction sites. Equipment installation and commissioning
require workers to have high welding operation and learning ability. The design integrates thecontrol system, welding machine, wire feeder, and other components to increase portability
and provide standardized welding theory and operation training for operators, enhancing
welding efficiency.

(a) a pile-beam integrated (b) welding process of bridge (c) the appearance of the weld(PBI) machine with apipe pile
welding robot
Fig. 3. Application of welding robot in integrated construction of bridge pile foundation
5. Conclusion
The paper presents a welding robot designed for splicing prestressed pipe piles, whose
reliability and technical advantages are demonstrated through indoor simulation welding tests
and bridge pipe pile applications. The main conclusions are as follows:(1) The welding robot adopts a puller type quick locking mechanism, which can realize thequick locking of the steel belt guide rail.
(2) The welding robot adopts the welding process of three layers of welding. The surface of
the weld seam is full and uniform and there is no internal defect, which can meet the welding
quality requirements.
(3) For prestressed pipe piles with a diameter of 1 m, it takes about 40 minutes to complete
the automatic welding of a single weld, which is 1/3 less than manual welding, and the labor
intensity of workers is significantly reduced.
6. References[1] Yang Zhijian. Research on Seismic Performance of Prestressed High-strength Concrete
Pipe Piles. Tianjin University, 2014.
[2] Tang, Mengxiong, Zao Ling, and Yuliang Qi. 2023. Bending Strength of Connection
Joints of Prestressed Reinforced Concrete Pipe Piles, Buildings 13, no. 1: 119.
https://doi.org/10.3390/buildings13010119.

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[3] Sun Xiaoli, Zhou Zhiguo, Wang Wenjian, Yang Jun, Hu Liangjun, Zhao Jinming, Yuan
Qibao, Zou Kai, Xie Huihua. Quick assembly and disassembly of pipe pile welding vehicle
with locking track [P] Patent No.: ZL 202110638939.0. 2022.03.11[4] Sun Xiaoli, Zhou Zhiguo, Wang Wenjian, Yang Jun, Zhao Jinming, Yuan Qibao, Xie
Huihua, Chen Jinhai. A quick docking and automatic welding equipment for pipe piles [P]Patent No.: ZL. 2020 2 1876358.8. 2021.05.24.

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Structural Behavior of Precast Concrete Beam-Column Joints
Jaehyung Heo1, DongIk Shin1, Yeon Je Choi1, Dong-Hwan Kim2, Hyeong-Gook Kim2,
Kil-Hee Kim2, and Jung-Yoon Lee3*
1: Department of Civil, Architectural, and Environmental System Engineering, SungKyunKwan
University, Suwon, Republic of Korea; email: gjwogud97@naver.com
2: Department of Green Smart Architectural Engineering & Urban Systems Engineering, Kongju
National University, Kongju, Republic of Korea; email: kimkh@kongju.ac.kr
3: School of Civil, Architectural Engineering and Landscape Architecture, SungKyunKwan
University, Suwon, Republic of Korea; email: jungyoon@skku.edu*: corresponding author

Keywords: precast concrete; beam-column joints; seismic design; energy dissipation; stiffness
Abstract: Precast concrete (PC) technology has been known for a long time, and the advantagesof PC structures have been widely known, but there are not many cases of constructing high-rise
PC structures. In order to construct high-rise PC structures, it is necessary to secure technologies
such as bonding, assembly, and lifting, and in particular, earthquake-resistant design is required.
The basic design principle of seismic design of PC structures is that PC structures should secure
the safety and serviceability required in monolithic structures. Different countries apply different
methods to assess the emulation of PC structures. For example, ACI 374.1-05, ACI PRC-550.109, and ACI ITG 5.1-07 evaluate the strength, energy dissipation and stiffness of PC structureson an absolute criteria. On the other hand, Japanese PC guidelines relatively evaluate thestructural performance of PC structures by comparing to the behavior of cast-in-place concrete
structures. In this paper, the structural performance of 160 specimens was analyzed to evaluate
the basic performance for the emulation of PC moment frame. The analytical results indicatedthat when the connection of PC members was properly designed, the structural performance of
PC members was almost similar to that of cast-in-place concrete members.
1. Introduction
The basic design principle of seismic design of precast concrete (PC) structures is that PC
structures should secure the safety and serviceability required in monolithic structures. It isdifficult to make the structural performance of PC structures equal to that of cast-in-place (CIP)
concrete. In the structural design codes, PC structures with performance above a certain limit
value are considered to have the same performance as CIP concrete structures. The design codes
certifies that the performance of a PC structure that exceeds a certain threshold is equal to that ofa CIP structure. These PC structures can be designed by applying seismic coefficients applied to
CIP structures. Performance evaluation of these PC structures is called emulation evaluation.

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Different emulation evaluation methods are being applied for PC structures in each country. For
example, in ACI 374.1-05 (2005), PC structures are considered to have equivalent performances
to CIP concrete structures when they exhibit the following behaviors:
1.

The test module shall have attained a lateral resistance (load or moment) equal to or
greater than En before its drift ratio exceeds the value consistent with the allowable story
drift limitation, where En is the nominal lateral resistance of test module determinedusing specified geometric properties of test members, specified yield strength ofreinforcement, specified compressive strength of concrete, a strain compatibility analysis
for flexural moment strength, and a strength reduction factor of 1.0

2. The maximum lateral resistance ( Emax ) shall have not exceeded exceed the lateralstrength ( λ En ), where λ is the specified overstrength factor for the test column.
3. The third complete hysteresis curve at the critical drift ratio shall have satisfied thefollowing conditions:- Peak force for a given loading direction shall have been not less than 0.75
Emax Emax for the same loading direction- The relative energy dissipation ratio shall have not less than 1/8.
- The secant stiffness from a drift ratio of –0.0035 to a drift ratio of +0.0035 shallhave been not less than 0.05 times the stiffness for the initial drift ratio.
These guidelines are the absolute emulation evaluation criteria. The absolute method evaluates
the performance of the precast structure based on its own characteristics, such as the strength,
stiffness, and load-bearing capacity. It does not compare the performance of PC structures withthose of CIP concrete structures.
On the other hand, “AIJ Guidelines for the Design of Structural PC Emulating Cast-In-place
Reinforced Concrete”(AIJ, 2002) evaluates emulation differently from the ACI 374.1-05
guidelines. In this guideline, the structural performance of PC structures must be verified bycomparing them with those of CIP concrete. That is, it is based on relative emulation evaluation
criteria, as follows (Lee et al., 2023):
1. The flexural yielding strength of PC members must be equal to or greater than that of the
CIP concrete members.
2. The strength of PC members must be equal to or greater than that of the CIP concrete
members.
3. The strength at the second complete hysteresis curve at the critical drift ratio must remain
within 80 %.
4. The deformation at flexural yielding of PC members should not differ by more than 20 %from that of CIP concrete members.
The relative method compares the behavior of the precast structure with that of an equivalentcast-in-situ (CIS) structure. This method is based on the assumption that the behavior of the CIS
structure is well understood and predictable. Both evaluation methods have their advantages anddisadvantages. The choice of method depends on the specific requirements of the structure andthe degree of accuracy desired. Evaluating emulation based on “absolute criteria”, such as in the283

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ACI 374.1-05 guideline, is simple, because the performance of CIP concrete members needs not
be reviewed. However, the relative energy dissipation ratio and stiffness are significantly
affected by the size and material properties of the member, leading to severe performancevariations. In general, the absolute method requires more detailed information about the structure
and the materials used. Conversely, emulation evaluation based on the relative criteria has theadvantage of faithfully reflecting the basic requirements for emulation. However, theperformance of CIP concrete members must be additionally evaluated. (Lee et al., 2023)
In this paper, the structural performance of 160 specimens was analyzed to evaluate the basic
performance for the emulation of PC moment frame.
2. Analysis of 160 Beam-Column Joints
To evaluate the emulation of PC structures, 160 beam-column joint specimens were collected in
the existing literature. Among the collected specimens, non-structural failure specimens, such as
damage to the connecting hardware or local failures, were excluded. Specimens from theliterature were tested for various variables. The 160 beam-column joints consisted of 45 CIP and
115 PC specimens. There were 133 internal joints and 27 external joints. To evaluate theemulation of the PC frame, the ductility ratio ( µθ = θu / θ y ) of the 160 beam-column joints wasevaluated according to the variables. The drift ratio ( θ y ) at flexural yielding was determined as
the drift ratio when 66.67% of the maximum load and the origin where the load was 0 wereconnected based on the proposal of Pan and Moehle (1989). The ultimate drift ratio ( θu ) is theratio when the load is reduced to 95 percent of the maximum load. Figure 1(a) shows theductility ratio ( µθ ) vs. the strength ratio ( V jn / V jby ) of the specimen. The strength ratio( V jn / V jby ) was calculated by dividing the joint strength by the beam flexural yield strength. The
specimens were divided into CIP and PC. In Fig. 1 (a), the ductility ratio ( µθ ) increases slightly
as the strength ratio ( V jn / V jby ) increases. When V jn / V jby is less than 1, µθ is approximately 2;
however, when V jn / V jby is 2, µθ increases to approximately 5. Specifically, θu exhibited analmost constant trend regardless of the type of concrete (CIP or PC). Figure 1(b) compares theductility ratios of the specimens classified by the ratio of the flexural strengths of the column andbeam ( ∑ M c / ∑ M g ). The relationship between ∑ M c / ∑ M g and ductility was similar to thatbetween V jn / V jby and ductility. That is, the ductility ratio ( µθ )increases slightly as∑ M c / ∑ M g increases. However, the effect was not as pronounced as that of V jn / V jby on µθ .

In Fig. 1(b), µθ also exhibits a constant trend regardless of the type of concrete (CIP or PC).
Consequently, because µθ shows an almost constant tendency regardless of the type of concrete,
it seems to satisfy the relative emulation that compares the performance of the CIP structure withthat of the PC structure. However, as µθ varies from approximately 0.7–9, showing a large
range of change, the evaluation result will be different depending on the member characteristics
if the emulation is evaluated based on absolute criteria. When comparing the strength of PC and
CIP specimens, the strength reduction rate of the two types of specimens was almost similar. The
strength of the specimen decreased as the drift ratio increased, so there were specimens that did
not satisfy absolute emulation evaluation criteria depending on the size and conditions of thespecimen.

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10

CIP
PC

8

8

6

6

µθ

µθ

10

4

4

2

2

0

0

1

2

3

V /V
jn

4

0

5

CIP
PC

0

1

2

3

4

M /M

jby

c

(a)

g

(b)

Fig. 1. Behavior of prestressing steel in strengthened girders subject to live load: (a) Ductility
ratio vs. V jn / V jby ; (b) Ductility ratio vs. ∑ M c / ∑ M g
3. Conclusions
The analytical results indicated that emulations of PC intermediate and special moment resisting
frames should be evaluated based on drift ratios of 2% and 3.5% respectively. The strength
reduction rate of PC member was almost the same as that of the CIP concrete member when thedrift ratio was 2 percent, but decreased as the drift ratio increased. The relative and absolute
emulation evaluation results were consistent for energy dissipation ratio and stiffness. However,
the absolute emulation evaluation method for special moment resistance frames was too strict for
special moment resistance frames.
4. Acknowledgments
This research was supported by the National Research Foundation of Korea (NRF) grant fundedby the Korea Government (MIST), under the project "Development of Seismic Design Method
for Reinforced Concrete Beam-Column Assembles of Moment Resistance Frames"(2023R1A2C3002443)
5. References
AIJ. 2002. Guidelines for the Design of Structural Precast Concrete Emulating Cast-in-place
Reinforced Concrete, Architectural Institute of Japan.
ACI T1.1-01. 2005, Acceptance criteria for moment frames based on structural testing and commentary, American Concrete Institute.
Lee, J.-Y., Heo, J., Byun, H.-W., Kim, K.-H., and Shin, D. 2023. Comparison of Two Emulation
Evaluation Methods of Precast Concrete Structures, Engineering structures, Submitted paper.
Pan, A., and Moehle. J.P. 1989. Lateral displacement ductility of reinforced concrete flat plates,
ACI Structural Journal, 86(3), 250-258.

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Diaphragm Wall Mono-Shaft as an Economical Time Saving Bridge Deep
Foundation
Yoni Pestes1* and Daniel Elmaleh212: Via Bridges design and engineering, Tel Aviv Israel; email: Yoni@via-bridges.com: Via Bridges design and engineering, Tel Aviv Israel; email: Daniel@via-bridges.com*: corresponding author

Keywords: Bridge foundation; Seismic design; Mono-shafts; Diaphragm wall boring
Abstract: Deep foundation is a widely used type of bridge foundation across the world.
Typically, round concrete piles are arranged in groups to support the bridge's columns andabutments. These columns can vary in type and shape, depending on the substructure'sfunctionality and aesthetic design. To connect the columns and the pile groups, a pile cap is used.
These underground structures must be of sufficient height to anchor the reinforcement from thebridge's columns and piles. However, reaching the piles' level often requires extensive
excavation, which can pose a challenge when building the bridge pier near existing structures ora roadway. One common solution to this issue is to use large circular mono-shafts. However, thisapproach results in a specific column type that is limited by the diameter of the shaft and theavailable space for the foundation's location. To offer an alternative foundation type, this paper
presents a case study of a diaphragm wall mono-shaft foundation. The study will outline theconsiderations that went into designing this foundation, taking into account relevant codes,
manufacturing, and construction limitations.
1. Introduction
The placement of bridge piers is a critical factor in determining the span length of a bridge
during the preliminary design phase. This decision can help identify the optimal bridge type andsolution and determine the super-structure's static height. By minimizing the need for excavationand getting as close as possible to the obstacle to be bridged, a more efficient and economicalsolution can be achieved. This paper focuses on the detailed design of the Pinhas-Rosen Bridge,
a light train bridge constructed in a dense urban area abundant with utilities and obstacles in themunicipality of Tel-Aviv, Israel. In this design-built (DB) tender, the client's bridge design in the
preliminary phase used conventional solution of column based on group of piles connected witha pile cap. This method required extensive excavation and construction of sheet pile wall when
getting close to the active road. The contractor's design team proposed an alternative mono-shaft
diaphragm wall solution that eliminated the need for expensive and time-consuming sheet pilewall and pile cap casting, resulting in a more economical and efficient solution. Additionally,
this approach reduced the maximum span length and optimized the static height of the pre-cast
prestressed U-shaped beams, helping the design team win the DB tender for the contractor, by
substantially reducing the cost of the project.

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2. Bridge Description
The Pinhas Rosen Bridge is located in the Tel-Aviv municipality of Israel and forms part of a
new 40km green line light train aimed at reducing traffic congestion in the center of Israel. The
bridge is designed to elevate the double-track green line over Pinhas Rosen Road, spanning alength of 321.8m and comprising 13 spans, with a maximum span length of 29.0m. The bridge isdivided into four sections ranging from 48.6m to 125.1m in length, each section is separated byexpansion joints (see Fig 1). The superstructure comprises of U-shaped precast prestressedgirders measuring 1.45m in height with a 0.25m cast-in-situ deck. The girders in the middle piersof each section are fixed to the column, while elastomeric bearings are used for the expansion
joint pier, allowing for an elastic connection. In the middle fix piers, the trapezoidal-shaped
concrete columns are directly placed and connected to the bridge foundation; concrete
rectangular diaphragm wall drilled mono-shaft foundation.

Fig. 1. Pinahs Rosen bridge longitudinal section
3. Rectangular Diaphragm Wall Drilled Mono-Shaft Foundation
The Pinhas Rosen Bridge incorporates several innovative design features that helped to reduce
the overall project budget and secure the tender. One main feature is the use of rectangulardiaphragm wall drilled mono-shaft foundations. This approach eliminated the need for an
oversized pile cap that would have required massive excavation. It should be noted, however,that the absence of pile caps meant that the boring process required meticulous attention to detail.
To minimize deviations during boring and construction, the contractor used concrete guides foreach shaft.Given that the bridge supports two rail tracks, each pier is composed of two columns
connected directly over the foundation at the ground level with a non-contact spliced rebar
connection.
The bridge's piers can be classified into two types. Piers 2-4 are configured with four U
precast beams and one deck, resulting in a pier in the form of a frame as can be seen in Fig 2. (a).
This configuration allows for a park walking and bicycle route to pass between the frame's
columns, resulting in a smaller column contour compared to the foundation shape (Type II shaft,Caltrans 2019). For this type of pier, a 1.2m x 2.8m diaphragm wall foundation was used. On theother hand, piers 5-13 support train tracks that are split, so each column is individually supported
between 2-3 U precast beams as can be seen in Fig 2. (b). This results in a larger column contour
compared to the foundation shape. For this type of pier, a 1.0m x 3.2m diaphragm wall287

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foundation was used. The foundation dimension was chosen to account for the pier's vertical load,
element capacity, and contractor equipment

(a)

(b)

Fig. 2. Pinahs Rosen bridge cross sections - (a) Piers 2 and 4 frame; (b) Pier 5-13 fixed column
4. Column and Diaphragm Wall Design
The bridge design was developed using SOFiSTiK FEM software, with the aim of accurately
modeling the bridge's geometry. The superstructure was modeled using the grillage method, anda rigid connection was established between the superstructure and the columns, taking into
account the diaphragm transverse beam casting between the longitudinal beams and the columns.
The columns were modeled using beam elements, which were directly connected to thediaphragm wall foundation that was also modeled as beam elements (refer to Fig. 3).

(a)

(b)

Fig. 3. Pinahs Rosen bridge FEM of the piers (a) Piers no. 2 frame; (b) Pier no. 12 column

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To model the interaction between the structure and the soil, springs were used in the X and Y
directions. The soil profile in the area of the bridge consisted of clayey sands and sand-clay
mixtures (SC) up to a depth of 5.0m, and sand with fines and gravel (SP-K1) below that. The
horizontal soil subgrade reaction was calculated to be 14,000kN/m3 between depths of (-1.0) (-5.0m) and 30,000kN/m3 below. The design of the column was done taking into consideration
vertical service loads due to dead load, temperature, creep and shrinkage, live loads such as train
movement and pedestrians. Also, horizontal service live loads from braking, traction andcentrifugal loads of the train. Additionally, the design considered extreme event loads such as
earthquake and wind loads. For earthquake design, modal analysis was carried out, which
considered the frequencies of the bridge and local ground acceleration to predict the horizontal
forces due to a seismic event.
The earthquake modal analysis, together with the calculation of horizontal forces
resulting from constant loads (e.g., concrete creep and shrinkage) generated a bending moment
diagram (refer to Fig. 5). Fig. 5(a) illustrates the bending moment diagram of pier 2, displaying aframe-like behavior in both directions, with the bending moment tending to zero at ground level,
making it an ideal location for splicing the rebar. Fig. 5(b) shows the bending moment diagramof pier 12, with a frame-like behavior in the longitudinal direction, where the bending moment
also tends to zero at ground level, providing an optimal location for splicing the rebar. In thetransverse (robust) direction of the column, the maximum bending moment occurs near the mid
depth of the foundation, far from the spliced area. As the transverse direction is robust, thecolumn's bending moment capacity is much larger than the calculated bending moment. Thissolution allows the bridge design to meet the Israeli standard (IS) and Caltrans requirements forseismic event.

(a)

(b)

Fig. 5. Bending moment for seismic loading combination and bending moment capacity inlongitudinal(y) and transverse(z) direction - (a) Pier 2; (b) Pier 12
5. Rebar Design
The calculations were performed considering bi-axial bending moments, as specified in EN1992-1-1-2004 section 5.8.9 and accounted for the increase in bending moment resulting fromconstruction imperfections and buckling. To determine the active length of the column forbuckling, the point of fixity method, as described by Robinson et al. (2006), was used. The
column's capacity was calculated using an interaction diagram, which takes into account the axial
loading and the appropriate bending moment capacity. It is important to note that the design ofthe column and the foundation top half included a minimum of 1.0% reinforcement inaccordance with IS 1227/1.

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The produced forces for design, including axial, shear, and bending moments, were compared tothe calculated column capacity. Based on this analysis, the column can be divided into four
critical parts.
a) The top of the column is connected rigidly to the superstructure and is anticipated to undergo
the greatest bending moment in the longitudinal weak direction and transverse robust direction,
making it the predicted location of the plastic hinge. In accordance with IS 1227/1, confinement
reinforcement needs to be provided for a section of 80% of the maximum bending moment or lc,
which is the section height in the direction of seismic force, whichever is greater.
b) The ground level, or splice zone, marks the connection between the diaphragm wall
foundation and the column. As this type of connection is not covered by Israeli standards,
AASHTO and Caltrans limitations were used. For piers 2-4, which have oversized foundations
compared to the columns (Type II shafts), an inner longitudinal rebar, smaller than thediaphragm foundation contour, was embedded 3.0m (Dc,max + 48lbl) according to Caltrans 2019section 8.3.2. The inner longitudinal rebar protrudes and is connected with a non-contact spliceto the column. This type of connection was studied by Masud et al., (2020) and can be seen inpiers 2-4 and 5-13. The rebar splice was embedded 2.0m into the column to be as close as
possible to the column longitudinal rebar. The distance between the foundation protruding rebar
and the column rebar was limited to 4ϕ(longitudinal rebar) or 50mm, as per EN 1992-1-1:2004
section 8.7.2. If this distance was exceeded due to the column's trapezoidal shape or contour, theoverlap was increased by the clear space between the longitudinal rebars. Additional transverse
reinforcement was added to fully develop the capacity of the longitudinal non-contact lap spliceand prevent brittle failure. It's important to note that the element's capacity in this section wasdetermined based on the smaller cross-section of the column (pier 2-4) or diaphragm foundation(pier 5-13).
c) The location of the bottom maximum bending moment was checked for bi-axial bending
moment and met the requirement of 1.25Mmax < Mcap (Caltrans 2019). This is in order to reduce
the potential for plastic hinge to form in a location that is difficult to examine.
d) The bottom half of the foundation was checked with the minimum required reinforcement in
the section (0.6% of the overall cross section) also making sure that 1.25Mmax < Mcap (Caltrans2019). limitation was fulfilled.
Overall, the design of the bridge column was thoroughly analyzed and optimized to ensure its
structural integrity under various loads and conditions.
6. Conclusion
This paper introduces a novel method for constructing a monolithic diaphragm wall bridgefoundation. The rebar configuration in the foundation and the column was carefully designed.
Examining the produced bending moment in both directions to predict the location of potential
plastic hinges. This examination together with construction limitation led to be preferred location
for the splice zone at the ground level. The method offers several advantages, such as faster
construction without the need for a pile cap, the ability to place the columns and foundation
closer to obstacles or roadways, and a greater variety of column shapes compared to circular

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mono shafts. However, there are also some potential disadvantages to consider. Firstly, the highamount of reinforcement needed for the column and diaphragm wall foundation can makethreading the diaphragm wall foundation into the column a challenging task. Secondly, it isessential to carefully investigate the location of potential plastic joints that may form during aseismic event and pay close attention to the location and details of the splice rebar. Lastly, it is
crucial to be aware that using a column on top of a single foundation element can increase theinfluence of deviation of the pier and the magnitude of the bending moment that can result from
eccentricity.
7. References
AASHTO. AASHTO LRFD Bridge Design Specifications, 8th ed. Washington, D.C.: American
Association of State Highway Transportation Officials; 2017.
Caltrans, Seismic Design Criteria Version 2.0, State of California Department of Transportation,
April 2019.
EN 1992-1-1 (2004) (English): Eurocode 2: Design of concrete structures - Part 1-1: Generalrules and rules for buildings [Authority: The European Union Per Regulation 305/2011,
Directive 98/34/EC, Directive 2004/18/EC]
Israeli standard 1227 part 1 – loads on bridges: highway bridges, 1988.
Masud, M., Chen, H., Sawab, J., Huang, H.W., Xu, B., Mo, Y.L. and Hsu, T.T., 2020.
Performance of non-contact lap splices in geometrically dissimilar bridge column to drilled shaft
connections. Engineering Structures, 209, p.110000.
Robinson, B., Suarez, V., Robalino, P., Kowalsky, M. and Gabr, M., 2006. Pile bent design
criteria. NCDOT Research Project 2005, 19.

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Performance Comparison of Nonlinear Dampers for Suppressing VortexInduced Vibration of Flexible Bluff Structures
Jingwei Zhang1, Xiaojun Wei1*, and Xuhui He1
School of Civil Engineering, Central South University, Changsha, China
Email: zhangjingwei2019@csu.edu.cn; xiaojun.wei@csu.edu.cn; xuhuihe@csu.edu.cn*: corresponding author

Keywords: Nonlinear dampers; Vortex-induced vibration; Flexible bluff structures; Multiple
modes
Abstract: Flexible bluff structures under wind excitations are prone to vortex-induced vibrations(VIVs). Linear dampers designed for a single mode may be ineffective in suppressing VIV whenthe excited mode is not the target mode. Nonlinear dampers, characterized by broadband
vibration energy absorption, are more appropriate for suppressing VIV involving multiple modes
and/or with the dominate mode varying with wind velocity. This paper aims to provide a
performance estimation of nonlinear dampers, with cubic stiffness element or/and cubic damping
element, in VIV mitigation of flexible structures. The governing equations of fluid-structuredamper systems are established, where the vortex-induced force is described using a wakeoscillator model in consideration of fluid-structure interaction. A flexible circular cylinder is
employed for the performance comparison of a linear tuned mass damper and three nonlinear
dampers in mitigating VIV involving multiple modes. The optimal parameters of all the dampers
are identified using a brute force approach. The research outcomes are helpful in developing asuitable nonlinear damper for VIV control of flexible bluff structures.
1. Introduction
Vortex-induced vibration (VIV) of flexible structures may result in serviceability and fatigue
problems (Wu et al., 2012). Adding dampers is one of the main countermeasures to control
undesired VIV (Fujino and Siringoringo, 2013). Among them, linear dampers are used widely
due to its simplicity. However, a linear damper usually mitigates the structural vibration
dominated by a single target mode only. Nonlinear dampers, featured by broadband vibration
control capability, are promising for VIV mitigation involving multiple modes and/or with thedominant mode varying with wind velocity. (Gendelman et al., 2001). Nevertheless, it is unclear
what type of nonlinear dampers is the most appropriate one. This paper aims to compare thecontrol performance of dampers with cubic damping or/and cubic stiffness for suppressing VIVof flexible bluff structures. A flexible circular cylinder with a linear/nonlinear damper is
employed for numerical demonstration purposes.
2. Governing Equations of Fluid-Structure-Damper Systems
This section describes the mathematical modelling of coupled fluid-structure-damper systems,
where different types of dampers are considered. The schematic of a coupled system is shown in292

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Fig. 1. The primary structure is assumed to be an elastically supported circular cylinder subjected
to uniform wind and vibrate only in the cross-flow direction. The vortex-induced force is
described through a wake oscillator model (Facchinetti et al., 2004). A damper is attached inside
the cylinder to mitigate VIV of the structure. Four types of dampers considered here, namely, a
linear tuned mass damper (LD), a single degree of freedom (SDoF) nonlinear damper with linear
viscous damping and cubic stiffness (ND1), a SDoF nonlinear damper with cubic viscousdamping and linear stiffness (ND2), and a SDoF nonlinear damper with cubic viscous damping
and cubic stiffness (ND3). The governing equations of a coupled system can be expressed inmodal space as

where ys = ∑ n=1 ϕ n ( x )ηn , q = ∑ n=1ϕn ( x)qn and yd are structural displacement, wake variable,NN

and displacement of the damper, respectively. N represents the total number of considered
modes. The over dot denotes the derivative with respect to time t . ηk and qk are the kth-ordermodal coordinates of ys and q , respectively. mk , ωk , ξk and ϕk ( x) are the modal mass, natural
frequency, damping ratio and mode shape of the kth-order mode of the primary structure,
respectively. x , xd and L denote the coordinate along the length direction of the structure, thelocation where the damper is installed and the total length of the structure, respectively. md , cd ,i ,kd , j and Fdi, j are the mass, damping, stiffness and the force of the damper, respectively. i and j
represent the power of the damping term and stiffness term, respectively. Fw is the VIV force
exerted on the structure, which is assumed to be fully correlated along the structural length. D ,U and ρ are the feature dimension of the structure, uniform wind velocity and air density( ρ = 1.225 kg/m3 in this paper), respectively. ω , γ , CL0 , ε and A are the vortex-shedding
frequency, fluid-added damping coefficient, reference lift coefficient, van der Pol parameter andscaling parameter, respectively.

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Fig. 1. Schematic diagram of a coupled fluid-structure-damper system.
3. Numerical Example
A flexible circular cylinder is considered as the primary structure in this section (Zhang et al.,2022). The cylinder is assumed to be a simply supported Euler-Bernoulli beam. The mode shape,
natural frequency and modal mass can be determined from 𝜑𝜑𝑘𝑘 (𝑥𝑥) = 𝑠𝑠𝑠𝑠𝑠𝑠(𝑘𝑘𝜋𝜋𝑥𝑥/𝐿𝐿),
𝐿𝐿𝜔𝜔𝑘𝑘 = 𝑘𝑘²𝜋𝜋²�𝐸𝐸𝐸𝐸/(𝑚𝑚�𝐿𝐿4 ) and 𝑚𝑚𝑘𝑘 = 𝑚𝑚
� ∫0 𝜑𝜑𝑘𝑘 2 (𝑥𝑥|)𝑑𝑑𝑥𝑥 , respectively. The first five modes are considered
to estimate the performance of different dampers on VIV control involving multiple modes, andthe damping ratio of each mode is set to be ξk = 0.13% . The physical mass per unit length, total
length, feature dimension and stiffness of the cylinder is taken to be m = 0.044 kg/m , L = 3 m ,D = 0.02 m and EI = 1.7815 Nm2 , respectively. Parameters in the wake oscillator model aretaken to be St = 0.2 , γ = 0.8 , CL0 = 0.3 , ε = 0.3 and A = 12 (Facchinetti et al., 2004). The
responses of the coupled systems are estimated using the Newmark-beta numerical integration
method with initial conditions of ηk (0) = 0.01D . Parameters of the damper installed at the mid-span of the cylinder are optimized through parametric analyses that aims to minimize thestructural steady-state displacement response. Both linear and nonlinear dampers are designed tominimize the 1st-mode VIV response at the wind velocity U = ω1D / 2π St . The mass ratio is set
to be µd = md / m1 = 5% . The optimal stiffness and damping coefficients of dampers are listed inTable 1.
Table 1. Optimal parameters of different dampers.
LD
ND1
ND2
ND3
a
1
5
1
−
−
Stiffness coefficient1.56 ×10
6.50 ×10
1.56 ×10
6.50 ×105
Damping coeffcientb
6.37
1.44×10−4
8.93×10−3
6.28×102a: units of linear, quadratic and cubic stiffness are (N/m), (N/m2) and (N/m3), respectively.
b: units of linear, quadratic and cubic damping are (Ns/m), (Ns2/m2) and (Ns3/m3), respectively.
Fig. 2 presents the performance of linear and nonlinear dampers on VIV control of the flexible
circular cylinder. The linear damper LD designed for 1st-mode can mitigate only the VIV of thistarget mode. Moreover, when the vortex shedding frequency is not equal to the 1st-mode natural
frequency, the control performance of the linear damper is inferior to that of the nonlinear
dampers. The nonlinear dampers ND1, ND2 and ND3 are capable of controlling VIVs involving

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multiple modes, e.g. 1st-mode, 3rd-mode and 5th-mode, due to its broadband vibration
absorption capability. ND2 exhibits similar performance in suppressing the 1st-mode and 3rdmode VIVs to that of ND1 but better performance in suppressing the 5th-mode VIV. This meansthat dampers with cubic viscous damping are more effective for suppressing higher-order mode
VIV than those with cubic stiffness. Compared with ND2, ND3 is capable of slightly narrowing
lock-in ranges of VIV due to the introduction of cubic stiffness, while ND3 has worse control
performance in controlling the 5th-mode VIV. Noted that nonlinear dampers are ineffective incontrolling 2nd-mode and 4th-mode VIVs. This is due to the fact that the damper is installed at
the mid-span, the node of these modes.

Fig. 2. Performance of different dampers on VIV control of the circular cylinder. (RMS(ys/D)
denotes the root-mean-square value of the dimensionless steady-state displacement of thecylinder at the location with the largest response.)
4. Conclusions
Nonlinear dampers are superior in VIV control of flexible bluff structures to the linear damper
due to their capability of mitigating VIV involving multiple modes and their insensitivity to thevariation of vortex-shedding frequency. Among them, dampers with cubic viscous damping
nonlinearity are more effective than those with cubic stiffness or with both cubic stiffness andcubic viscous damping.
5. References
Facchinetti, M. L., De Langre, E. and Biolley, F. 2004. Coupling of structure and wakeoscillators in vortex-induced vibrations. Journal of Fluids and Structures, 19, 123-140.
Fujino, Y. and Siringoringo, D. 2013. Vibration mechanisms and controls of long-span bridges: areview. Structural Engineering International, 23, 248-268.
Gendelman, O., Manevitch, L. I., Vakakis, A. F. and M'Closkey, R. 2001. Energy pumping in
nonlinear mechanical oscillators: Part I - Dynamics of the underlying Hamiltonian systems.
Journal of Applied Mechanics-Transactions of the ASME, 68, 34-41.
Wu, X., Ge, F. and Hong, Y. 2012. A review of recent studies on vortex-induced vibrations oflong slender cylinders. Journal of Fluids and Structures, 28, 292-308.
Zhang, M. J., Wu, T. and Oiseth, O. 2022. Vortex-induced vibration control of a flexible circularcylinder using a nonlinear energy sink. Journal of Wind Engineering and Industrial
Aerodynamics, 229.

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Control the Torsional Flutter of a Rectangular Plate with Flexible Membraneat Its Leading Edge
Ziqiang Zhang1, Hanfeng Wang1, 2*, Zhiwei Liu1, and Chi Zhang1
1: School of Civil Engineering, Central South University, Changsha, China: National Engineering Research Center for High-speed Railway Construction, Changsha,
China; email: wanghf@csu.edu.cn

2*: corresponding author

Keywords: Torsional flutter; Leading edge vortex; Serrated flexible membrane; Flow control
Abstract: The wind resistance performance of large span bridges in mountain canyons is a key
concern in current wind engineering. Due to the complex wind environment of canyon terrain, theaverage wind attack angle (α) of bridge deck can often reach ± 5°, obviously larger than the rangeof ±3° in the criterion. Traditional passive aerodynamic measures (such as wind nozzles, etc.) may
be difficult to meet the complex wind resistance requirements of large span bridges in mountainous
areas. Therefore, a novel adaptive deformation passive aerodynamic measure is proposed. In thispaper, a simplified rectangular plate with a width to thickness (W/D) ratio of 20, which had one
degree of freedom in torsional direction, was used as the tested model. Torsional circular frequencyof the model = 40.9 rad/s, mass per unit length m = 1.67 kg/m and damping ratio ζ = 1%. Serrated
flexible membrane, with a Young's modulus of 0.2 Gpa, a thickness of 0.06 mm, and a length of
2D, were fixed at the leading edge of the plate to test their effects on its torsional flutter. All
experiments were conducted in an open-circuit low-speed wind tunnel with a test section of 450
mm in width, 450 mm in height, and 1200 mm in length. The tested α was 5 °, which wasconsidered as the most unfavorable angle of attack. The torsional vibration amplitude of the plate
was measured, with the reduced wind speed Ur ranged from 0 - 8.85, and the corresponding
Reynolds number (Re) was 0 - 1.55◊105. The flow field around the plate was also measured using
Particle image velocimetry (PIV) to reveal the control mechanism. For the uncontrolled case, theflutter starts at Ur = 6.77. For the controlled case with the membrane utilized, the membrane starts
flapping at Ur = 6.52, prior to the flutter of the plate. Once the membrane flaps, it significantly
suppresses the flow separation from the leading edge of plate, reducing the separation bubble on
its upper surface. The leading-edge vortex (LEV), which directly associated with the torsional
flutter, is also considerably weakened by the flapping membrane, together with thus reducing thetorsional moment of the plate. The experimental results shows that the flutter critical wind speed
increases by approximately 19 %, from 6.77 to 8.07, when the flexible membrane is utilized.
1. Introduction
With the construction of large span bridges across the sea and mountainous areas, the windenvironment at the bridge site has become increasingly complex. According to observation
analysis and wind tunnel simulation research on strong wind characteristics in coastal andmountainous areas (Pang, 2006), the maximum wind attack angle of bridge deck height is296

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generally greater than 5 °, and the flutter performance of bridge section will be further weakened
at high attack angle. However, previous study (Zhao et al., 2015) have shown that traditional
aerodynamic measures do not have universal control for complex inflow and wind-induced
vibration. Inspired by adaptive flow control measures in aerospace field (Tan et al., 2021; Gong etal., 2021; He et al 2022;), a new type of lightweight, flexible and deformable passive aerodynamic
measure is designed.
2. Experiment Details
All experiments were conducted in an open-loop low-speed wind tunnel with a test section of
0.45m in width, 3m in height, and 1.2m in length, respectively. More details of the wind tunnel
are introduced in Liu et al. (2023). The test model was made of abs material, with length L =420mm, width B = 200mm and thickness D = 10mm, as shown in Fig.1a. Serrated flexible
membrane with q = 10mm, s = 10mm and h = 5 mm was installed on the leading edge of therectangular plate at 5° wind angle of attack, as shown in Fig.1b. The main parameters of the modeland flexible membrane are shown in Table 1:

Serrated flexible membrane
y

U→

x
D
Pitch axis
s
q

h
2D
L

(a)

(b)

Fig. 1. Experimental Setup: (a) 3D Sketch; (b) Cross-section of the model and shape of serrated
flexible membrane
Table 1. Main parameters of rectangular plate model of serrated flexible membrane
Main parameters
Damping ratio /%1.00
Natural circular frequency /(rad/s)
40.90
Natural frequency fd /Hz
6.51
System stiffness /(Nm/rad)9.37
2
Rotational inertia /(kg.m )0.0056
Material of membrane
polyethylene
Membrane thickness /mm
0.06
Young's modulus of membrane /Gpa
0.20

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3. Results and Discussion
3.1 Amplitude characteristic
In fig.2, it can be found that nonlinear soft flutter occurs on the rectangular plat of both cases,
consistent with the phenomenon reported in Zhu et al 2015. The rectangular plate remains a stable
limit cycle vibration after the critical wind speed for vibration starting, and the vibration amplitude
increases approximately linearly with the increase of wind speed as shown in fig.3. In theuncontrolled case, the torsional flutter of rectangular plate occurs at Ur = 6.77. After installing
flexible membrane on the plate, the critical flutter speed of rectangular plate increased to about
8.07, which is about 19% higher than the uncontrolled case.
Uncontrolled caseControlled case20

Ur = 7.8

Ur = 7.8

12
Tortional angle (°)

Maximum tortional angle (°)

24

16
12
8

8
4
0
-4
-8

4

-12

0
0

2

4

6

8

10

Reduced wind velocity ( Ur=U/(fd B) )Fig. 2. Variation of with Ur

0

2

4

6

8

10

12

14

16

18

20

T/s

Fig. 3. Displacement time history with Ur =7.8

3.2 Flow characteristics
Flow characteristics of flow field above the fixed rectangular plate at 5° and Ur = 7.80 wasinvestigated by PIV. The maximum torsional amplitude of uncontrolled case is 8°, while thecontrolled case remains almost stationary at Ur = 7.80. Fig.4 shows the time averaged vorticity
filed and streamwise velocity which is calculated using the 2000 random sampled PIV snapshots.
The flapping configurations of the flexible membrane was recorded using a high speed camera by
Liu et al 2023 as shown in Fig.4b. In the uncontrolled case, there is a large-scale leading-edge
vortex above the rectangular plate as shown in Fig.4a. It can also be observed that the flow
separates from the leading edge of the rectangular plate and reattaches to the near trailing edge ofthe plate, coupling with the wake in Fig.4c. The initiation, separation and reattachment of leadingedge vortex creates a large separation bubble, however, a large mount of airflow on the uppersurface of the rectangular plate separates, resulting in a sharping change in the pitching momentof plate, which may lead to aerodynamic negative damping and torsional flutter.
In the controlled case, it can be discovered that the sizes of leading-edge vortices andseparation bubble above the rectangular plate were both significantly reduced by the flapping
serrate flexible membrane in Fig.4b,d. The reattachment point of flow moved forward from thetrailing edge to the position of 0.6B of rectangular plate and the vortex core also moved forwardfrom 0.55B to 0.22B of rectangular plate, compared to uncontrolled case. Consequently, theflapping serrated flexible membrane can weaken the rapidly changing in pitching moment causedby separation of airflow, thereby delaying the occurrence of torsional flutter.

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(a)

(b)

(c)

(d)

Fig. 4. Mean spanwise vorticity (top) and mean streamwise velocity (bottom): (a) (c) for
uncontrolled case; (b) (d) for controlled case
4. Conclusion(1) The new type of lightweight, flexible and deformable passive aerodynamic measure is designed
and investigated, which is expected to provide a better option for bridge aerodynamic measures in
the future.
(2) The flapping serrated flexible membrane can increase the flutter critical wind speed ofrectangular plate by approximately 19%, from 6.77 to 8.07.
5. Reference
Zhao Lin, Ge Yaojun, Guo Zengwei and Li Ke, 2015 Review and Reflection on Wind Vibration
Control of Long Span Cable Supported Bridges - Passive Control Effects and Active Control
Strategies for Main Beams Journal of Civil Engineering, 48 (12), pp. 91-100
Gong Xu'an, Zhang Xin, Ma Xingyu, Fan Ziye and Jiang Nan, 2021 Bionics overlaid control airfoil
flow separation experiment Journal of Aerodynamics, 39 (6), pp. 184-195
Tan, J., Wang, Z. and Gursul, I., 2021. Self-excited flag vibrations produce post-stall flow control.
Physical Review Fluids, 6(10), p.L102701.

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Liu, Z., Wang, H. and Zhao, C., 2023. Flow over a forward-facing step with a flexible membraneat its leading edge. Experimental Thermal and Fluid Science, p.110922.

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Modeling and Advanced Analysis

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Unseating of Bridge Decks-Codal Provisions and Fragility Analysis
Shambhavi Dube1*, Nirav Thakkar2, and Durgesh C. Rai3
1

: Doctoral Student, Dept. of Civil Engineering, Indian Institute of Technology Kanpur, Kanpur
208016, India; email: shamdube@iitk.ac.in2: Frmr. Graduate Student, Dept. of Civil Engineering, Indian Institute of Technology Kanpur,
Kanpur 208016, India; email: nirav.thakkar@gmail.com
3: Professor, Dept. of Civil Engineering, Indian Institute of Technology Kanpur, Kanpur 208016,
India; email: dcrai@iitk.ac.in*: corresponding author

Keywords: bridge bearing; elastomeric pad; fragility analysis; seismic forces; unseating;incremental dynamic analysis (IDA)
Abstract: Bridges are lifeline structures, and their continued operation during and after
earthquakes is critical. In the last few decades, there has been a dramatic increase in bridge
failures caused by earthquakes worldwide. The use of elastomeric pads helps in reducing seismic
forces on the substructure. However, it generates higher displacement demand, causing problems
like unseating and collapse of the deck slab. The performance of a bridge under an earthquake
largely depends on the dynamic characteristics of the bridge which in turn depends on deck spanand pier height. To understand the effect of the elastomeric pad on the seismic performance of awide range of RC bridges, a group of nine bridges with various spans and pier heights wasconsidered. Primary components of the bridge i.e., precast concrete girders, elastomeric pads,
and piers were designed for these span lengths and pier heights as per Indian Codal provisions.
The bridges were modeled using the finite element software, SAP2000. The vulnerability of thebridges was investigated for three damage states, namely unseating, collapse, and low pressure.
The problem of instability and unseating is a function of the size of the bearing, the width of the
supports, and the minimum normal pressure on the bearings. Incremental Dynamic Analysis(IDA) of the sample bridges was carried out with twenty far-field ground motions recorded on
soft rocks or stiff soils, and displacement of bearing is recorded along all three directions, which
are further used for vulnerability analysis. Codal provisions for the requirement of a minimum
support length were reviewed and compared for various international codes. Fragility curves
have been developed corresponding to 60%, 80%, 100%, 120%, and 140% of the capacity for
RC bridges with elastomeric pad bearings for the instability, unseating, and low-pressure damage
states. A significant effect of the size of the elastomeric pad and seating width was observed on
fragility curves. Results for maximum acceptable PGA and fragility curves can be used todetermine the potential damages from earthquakes and to assign prioritization for retrofitting theexisting bridges.

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1. Introduction
Elastomeric pads, also known as neoprene pads, have been found to impose less seismic force onpiers compared to other types of bearings (Pan et al.,2010; DesRoches et al., 2004; Chiang,
2008). They are cost-effective, durable, simple to detail, fabricate, and install, while also
allowing for thermal expansion and girder rotation. Due to these advantages, elastomeric pads
are a popular choice for bridge bearings in developing countries like India. RC bridge decks
supported by elastomeric pad bearings can move freely over the substructure during anearthquake, enabling large displacements. However, an additional anchorage system may benecessary to handle excessive displacement demand (Chiang, 2008). In India, some seismic
codes in practice do not allow the use of elastomeric pads in high seismic zones, while others
permit their use without anchorage, restraints, or shear keys, even in highly seismic areas. Often,
bridge deck girders are not anchored to bearing pedestals or pier cap beams, making it likely thatthe deck will collapse due to excessive displacement during a design-level earthquake. Recent
earthquakes have shown that bridges with elastomeric bearing pads are prone to unseating andcollapse (Rai et al., 2006). These problems can be prevented by ensuring sufficient seat length,
as specified by international codes based on factors such as span length, pier height, groundacceleration, and seismic displacement demand. To evaluate the impact of elastomeric pads on
the seismic performance of RC bridges, a group of nine bridges with different spans (10 m, 20 m,
and 30 m) and pier heights (5 m, 15 m, and 25 m) was examined. The study defined damage
states based on deformations in the elastomeric bearings in both longitudinal and transverse
directions and the minimum normal pressure on the bearings.
2. Design of Primary Components of Bridges
A three-span simply supported concrete bridge is used for the present study as shown in Fig. 1a.
The seismic design is carried out in both longitudinal and transverse directions of the bridge. The
design forces are obtained by combining i) 100% longitudinal + 30% transverse and ii) 30%
longitudinal +100% transverse direction loading. Foundations and abutments are consideredrigid enough to consider that design ground accelerations are directly applied to the bottom of
the pier and the bottom of the abutment bearing. Nine bridge samples are considered with span
lengths of 10 m (short), 20 m (intermediate), and 30 m (long-span bridge); and pier heights taken
as 5 m (short), 15 m (intermediate), and 25 m (long period bridges). Type I, Type IV, and TypeVI AASHTO I-Beams are used for 10 m, 20 m, and 30 m spans.

(b)

(a)
Fig. 1. (a) SAP 2000 model of a three-span bridge; (b) Schematic elevation of bridge

2.1. Design of Elastomeric Pads
Elastomeric pads shall be designed to resist the shear stress generated primarily by axial
compression, lateral displacement, and girder rotation while also meeting the stability andserviceability criteria. The design load for bearings is calculated by combining dead loads andvehicle loads as per IRC 6 guidelines. The elastomeric pads were designed based on IRC 83(Part II) and the key specifications are provided in Table 1.

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Table 1. Design summary of elastomeric pads
Description
Design loads per bearing
Dead load (kN)
Imposed load w/o impact (kN)Horizontal load (kN)
Rotation of girder end (rad)Geometrical details
Overall length of pad (mm)Overall width of pad (mm)Side cover (mm)Thickness of internal layers (mm)Thickness of external layers(mm)Thickness of laminates (mm)Numbers of internal layers
Total thickness of pad(mm)
Total elastomer thickness (mm)10 m

Span length
20 m
30 m

90
345
25
2.0E-4

265
470
470
530
25
25
4.4E-4 11.7E-4

320
200
6
8
4
3
2
33
24

400
250
6
10
5
3
3
52
40

500
250
6
10
5
3
3
52
40

3. Finite Element Modelling
The bridge is modeled using CSI SAP2000. The modulus of elasticity for M30 grade concrete isdetermined to be 27.4 GPa based on IS 456:2000. The piers are modeled as elastic beam-column
elements, which are expected to be subjected to significant axial forces and bending moments
and are likely to experience plastic hinges. To account for this, P-M2-M3 hinges are modeled at
the top and bottom of the piers, following the guidelines in FEMA 356 to reflect the frame action
along the transverse direction of the bridge. The pier cap beam is modeled as an elastic beamcolumn element, while the bridge deck is modeled as a single equivalent elastic beam-column
element. The foundation is assumed to be rigid and provides support to the pier, restraining all
six degrees of freedom at the bottom of the pier. The abutment is rigid and elastomeric pads aremodeled as Friction Isolator link elements with a coefficient of friction of 0.2.
4. Provisions for Seating Width in Various National Codes
The recent version of the Indian Code for Seismic Design of Road Bridges IRC SP:114 prohibits
the use of elastomeric bearings in high seismic zones IV and V. Instead, PoT, POT Cum PTFE,
and spherical bearings must be used. Earlier codes lacked this restriction, leading to many
bridges with elastomeric bearings in these high seismic zones. Deck unseating occurs when the
maximum bearing displacement along the longitudinal direction of the bridge exceeds theminimum seating width requirement shown in Table 2 calculated using various national codes.

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Table 2. Comparison of required minimum support length as per various codes
Span Length, m
10
20
30
Pier Height, m
5
15
25
5
15
25
5
15
25
Period, second
0.77 1.26 1.56 1.20 1.76 1.98 1.33 1.85 2.28
Required IS 1893/IRC: 6 380 480 580 405 505 605 430 530 630
AASHTO380 480 580 405 505 605 430 530 630
minimum
CALTRANS 252 552 852 252 552 852 252 552 852
supportlength, mm Japanese code 750 750 750 800 800 800 850 850 850
As per IRC 6, the combined effects of dead and vehicle loads result in an elastomeric pad lengthof 320 mm, 400 mm, and 500 mm for spans of 10 m, 20 m, and 30 m respectively. Existingbridges with these pad lengths are prone to unseating failure during earthquakes, particularly
those with greater pier heights. Consequently, a vulnerability analysis should be conducted toassess the potential for unseating failure and to propose appropriate retrofitting measures.
5. Assessment of Fragility
Nine bridge configurations were designed for a 0.36g PGA. Figs. 2a and 2b compare the IS 1893
design spectrum to the mean spectrum of 20 selected ground motions provided in Thakkar, 2011,
normalized to the 1.0g hazard level. The Spectral Displacement Response Spectra (Fig. 2b) wereobtained through Incremental Dynamic Analysis (IDA) for the selected ground motions.
Newmark's method with γ = 0.5 and β = 0.25 is used for time integration and 5% proportional
damping is assumed for periods of 0.35 s and 2.5 s. The maximum displacement at bearings andminimum pressure on the elastomeric pad is obtained by performing IDA and further utilized tocreate fragility curves for three damage states under various hazard levels. Bearing instability
occurs when the maximum transverse displacement exceeds half the width of the bearing anddeck unseating occurs when the maximum longitudinal displacement exceeds the minimum
seating width (SE) requirements. A damage state of low pressure is defined as a minimum
normal pressure on the elastomeric bearing below 1 MPa for all 9 bridges in the sample.

(b)

(a)

Fig. 2. (a) Acceleration response spectrum; (b) Displacement response spectra
A fragility curve describes the probability of exceeding a damage state as a function of chosen
ground motion intensity parameter. The probability that the demand exceeds the structural
capacity can be computed using Eq. 1. whereare structural demand, structural capacity,
structural normal distribution functions, respectively; andare the logarithmic standard
deviation for demand and capacity, respectively. This study utilized the procedure described by
Pan et al., 2007 to generate fragility curves for bridges.

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(1)
6. Results and Discussion
The Figs. 3a-3c, concludes that with the same pier height, the probability of failure decreases
with an increase in the span of the bridge mainly due to increased weight distribution and hence
stabilizing pressure on the bearings. However, a noticeable increase in the probability of failure
occurs when the pier height increases from 5 to 15 meters for a constant span length. The
probability of failure remains relatively constant for further increase in pier height to 25 m likely
due to the dynamic properties of the bridges. From Figs. 3d-3f, it is observed that the risk offailure is highest for the low pressure damage state and lowest for the unseating damage state forany span length or pier height. This indicates that during an earthquake, there is a higher
probability of the normal pressure on elastomeric bearings dropping below the low pressuredamage threshold of 1 MPa. Transverse ground motion components can cause uplift reactions onsome bearings that are comparable with the reactions from gravity loads.

(a)

(b)

(c)

(d)

(e)

(f)

Fig. 3. Fragility curves for unseating of bridges with (a) 5 m; (b) 15 m; (c) 25 m pier height for
unseating damage state; comparison of damage states of 15 m pier height and (d) 10 m; (e) 20 m;(f) 30 m span length
7. Conclusion
Past earthquakes have taught us the importance of providing adequate seat length to prevent
bridge deck collapse. In general, the risk of failure is high for unseating damage state, even for
low hazard levels. Results showed that the risk of failure increases with pier height and decreases
with bridge span. Bridges with small piers have a low risk of unseating, indicating thatelastomeric pads may be suitable. The acceptable maximum PGA that can be allowed for the
sample bridges with elastomeric pad bearings was found in the range of 0.11g to 0.41g and theexisting bridges with a design PGA greater than 0.41g are susceptible to unseating failure andneed retrofitting. The fragility curves developed in this study can aid in assessing earthquakerelated potential losses and prioritizing retrofitting efforts such as seismic links, STUs, verticalhold-down devices, and anti-dislodgement devices in the horizontal direction. However, these306

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curves are based on approximations and a limited sample of bridges, so their accuracy andapplicability in practical situations need to be further investigated with a larger sample of bridges
with different span lengths and pier heights.
8. References
AASHTO. 2017. AASHTO LRFD Bridge Design Specifications, American Association of State
Highway and Transportation officials (AASHTO), Eighth edition, Washington DC.
BIS IS 1893-Part 3. 2014. Criteria for Earthquake Resistant Design of Structures - Bridges andRetaining Walls, Bureau of Indian Standards (BIS), New Delhi, India.
BIS / IS 456. 2000. Plain and Reinforced Concrete – Code of Practice, Fourth Revision, Bureau
of Indian Standards (BIS), New Delhi, India.
CALTRANS. 2019. Seismic Design Criteria, California Department of Transportation(CALTRANS), V2.0.
Chiang J C. 2008. Application of Functional Bearing System to the Seismic Retrofit of Bridge,
Fourth US-Taiwan Bridge Engineering Workshop, Paper 16, Princeton, New Jersey.
DesRoches, R., Choi, E., Leon, R.T., Pfeifer, T.A. 2004. Seismic Response of Multiple Span
Steel Bridges in Central and Southeastern United States. II: Retrofitted, Journal of Bridge
Engineering, ASCE, Vol. 9, No. 5, pp. 473-479.
FEMA 356. 2000. Pre-standard and Commentary for the Seismic Rehabilitation of Buildings,
Federal Emergency Management Agency, Washington, D.C. 2000.
IRC 83-Part 2. 1987. Standard Specifications and Code of Practice for Road Bridges Section: IX
Bearings - Elastomeric Bearings, Indian Road Congress (IRC), New Delhi, India.
IRC 6. 2017. Standard Specifications and Code of Practice for Road Bridges, Section: II Loads
and Load Combinations, Indian Road Congress (IRC), Seventh Revision, New Delhi, India.
IRC SP:114. 2018. Guidelines for seismic design of road bridges, Indian Roads Congress (IRC),
New Delhi, India.
Japanese JRA. 2012. Design Specifications for Highway Bridges: Part V - Seismic Design, Japan
Road Association.
Pan, Y., Agrawal, A. K., and Ghosn, M. 2007. Seismic Fragility of Continuous Steel Highway
Bridges in New York State, Journal of Bridge Engineering, ASCE, Vol. 12, No. 6, pp. 689-699.
Pan, Y., Agrawal, A.K., Ghosan, M., Alampalli, S. 2010. Seismic Fragility of Multispan Simply
Supported Steel Highway Bridges in New York State. II: Fragility Analysis, Fragility Curves,and Fragility Surfaces, Journal of Bridge Engineering, ASCE, Vol. 15, No. 5, pp. 462-472.

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Rai, D.C., Murty, C.V.R., Jain, S.K., Kaushik, H.B., Mondal, G., Dash, S.R., Tang, A.,
Yashinsky, M., and Eskijian, M. 2006. The Effect of the December 2004 Great Sumatra
Earthquake and Indian Ocean Tsunami on Transportation Systems in India’s Andaman andNicobar Islands, Earthquake Spectra, Vol. 22, No. S3, pp. S561-S579.
Thakkar, N. 2011. Performance evaluation of Elastomeric Pads as Bridge Bearings under
Earthquake Loads, M. Tech. Thesis, Department of Civil Engineering, IIT Kanpur.

308

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Finite Element Modeling of Compressive Membrane Action in Transversely
Prestressed Deck Slabs
Amir Sana
University of Wollongong in Dubai, Dubai, United Arab Emirates; sanaamir@uowdubai.ac.ae,
sanaamir.1919@gmail.com
Keywords: concrete; bridges; finite element; compressive membrane action
Abstract: It has been established in literature that compressive membrane action enhances thepunching shear capacity of a laterally restrained slab. In this research an attempt has been madeto compute the compressive membrane forces and to determine the level of the membrane action
developed in a laterally restrained and transversely prestressed concrete deck slab. It is difficult,if not impossible, to theoretically predict the membrane action that can be developed in a slab.
This issue can be resolved by adopting a finite element approach rather than using cumbersome
analytical methods. A 3D solid model of an existing bridge was developed in TNO DIANA 9.4.4
software package and several nonlinear analyses were carried out. The pre-/post processor used
for the finite element model in this research is midas FX+ (3.1.0) for DIANA. This paper
describes the step-by-step procedure to model compressive membrane action in the deck slab.
Such a numerical study will encourage engineers to use finite element as a tool to analyze
existing and new structures.
1. Introduction
In the Netherlands, many old bridges with thin transversely post-tensioned concrete decks castin-situ between the flanges of long, precast concrete girders, were found to be critical inpunching shear when assessed by the Eurocode 2 “Concrete structures” (EN 1992-1-1:2005,2005). However, no visible signs of distress are observed which may be due to the presence ofresidual bearing capacity (Amir, 2014). Bridges are usually designed on traditional flexural
theories and it has been discovered that under concentrated wheel loads, the laterally restrained
deck slabs mostly fail in punching shear rather than in flexure and at loads much higher than
expected due to “Compressive Membrane Action” (CMA) that occurs in slabs whose edges arelaterally restrained by stiff boundary elements. When a load is applied, the slab edges tend tomove outwards and the restraint induces in-plane arching forces in the slabs enhancing thebearing capacity. This, in combination with transverse prestressing, can significantly enhance thebearing (punching shear) capacity (Amir et al. 2016). One of the aims of this research was toprove that sufficient CMA can develop in the bridge deck slab enhancing its bearing capacity.
For this purpose, a 1:2 scaled model of a real approach bridge (Fig. 1) experimental, numericaland theoretical approaches were used to estimate the bearing capacity and CMA. This paper
describes the numerical modeling procedure to simulate CMA in the deck slab. The details of thematerial properties, test setup, loading pattern etc. can be found in (Amir, 2014; Amir et al.
2016).

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Fig. 1. Top view and front view of the experimental setup
2. Finite Element Modeling
For the nonlinear analysis of the deck slab, a 3D solid finite element model (Fig. 2) of the
prototype bridge deck was developed in TNO DIANA 9.4.4 (Midas FX+ 3.1.0, 2012). A finemesh of elements was provided around the loaded area and ducts were modeled only in that partof the slab. Away from the loaded area, coarse mesh of elements was provided to reduce thesimulation time. For concrete, an elastic-perfectly plastic model (CONSTA) was used incompression, an exponential softening curve (HORDIJK) was used in tension, and a “Total
strain crack rotating model” was used for smeared cracking. A fracture energy of 0.15 N/mm
was assumed for the deck slab concrete and the poisson ratio for all the concrete components,
was taken as 0.2. For the embedded grid reinforcement, the von Mises plasticity criterion wasused with a poisson ratio of 0.3. An incremental-iterative procedure was used for the nonlinear
analysis and the modified Newton Raphson method was used for the solution. The prestressing
load was applied to the bridge deck as an external pressure. The applied load was displacementcontrolled; therefore, a default force and energy based convergence criterion was employed.
More detail can be found in Amir et al. 2019.

(a)

(b)

Fig. 2. The finite element solid bridge model: (a) 3D view; (b) Transverse cross-section.

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3. Modeling of Compressive Membrane Action (CMA)
A new technique of providing a layer of composed elements was used in the model to calculate
the in-plane forces arising from CMA. DIANA 9.4.4 specifies a composed element via a baseelement and a composition of regular solid elements (Fig. 3). The base element layer forms areference plane for which the generalized moments and forces are calculated by integration in thelocal z direction over the elements in the composition of each base element. The selection of thetype of base element depended on the type of the solid elements in the composition. The main
application of composed elements is in the post processing of analysis results. They have no
mechanical properties of their own.
z78Composition
Base element

Reference plane

(a)

1

1
2

4

2

3
5

6
8

7
3

4

(b)

6
9

5

(c)

Fig. 3. Composed elements (Midas FX+ 3.1.0, 2012): (a) Base elements in a structuralcomposition; (b) CQ8CM base element for CHX60 brick; (c) Base element layer provided within
the solid element composition surrounding the loaded area in a typical slab.
The solid elements in the loaded area were CHX60 bricks, so for composed elements CQ8CM
was used (Midas FX+ 3.1.0, 2012) as shown in Fig. 3(b). The CQ8CM element is an 8 node,quadrilateral curved base element and can only be provided where CHX60 elements are used in acomposition. The default integration scheme is 3×3 which matches that of CHX60 element. The
thickness of the base element CQ8CM was given equal to the thickness of the structuralcomposition, i.e. the thickness of the deck slab (100 mm). Fig. 4 shows how the composed
elements were provided in the finite element bridge model. Since these elements were used forthe calculation of compressive membrane forces, they were only applied in the area of interest i.e.
the loading area and its surroundings minus the area having ducts because composed elements
can only be provided in structured meshes. It is worth noting that the in-plane forces obtained
were expected to include both the transverse prestressing and membrane forces.

(a)(b)
Fig. 4. Composed elements provided in the bridge model fine mesh strip: (a) 3D view of thecomposed element layer provided within the fine mesh area of the deck slab; (b) 3D wireframing view of the composed elements in the fine mesh slab strips extracted from the bridge
deck model.

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4. Results and Conclusions
Several nonlinear analyses were carried out varying the Transverse Prestressing Level (TPL).
Fig. 5(a) shows the relationship between the distributed in-plane force (compressive in nature)and the failure load for various TPLs at the ultimate stage. It can be observed that the in-plane
force increases with the increasing prestress and the relationship is almost linear. Subtracting theinitial prestressing force from the total in-plane force corresponding to that particular TPL gives
a constant value of the compressive membrane force (CMF ≈ 370 N/mm) for the deck slab
showing that CMF alone is independent of the transverse prestressing level. This is also seen in
in Fig. 5(b) that shows the relationship between TPL and the in-plane force (sum of thetransverse prestressing force and the compressive membrane force) developed in the bridge deck.
It can be concluded that regardless of the magnitude of the prestressing applied, a certain level ofdefault compressive membrane force is developed in the plane of the deck slab due to the built-in
restraint available in the form of edge supports (girders), diaphragms, surrounding slab area etc.,
if all other parameters remain the same (type of loading, concrete strength etc.).

Punching failure load [kN]375

~ 370 N/mm

CMF
In-plane= CMF + TPL
Linear (In-plane= CMF + TPL)

350
R² = 0.9995

325
300
275

501 N/mm
423 N/mm 1.25 MPa

250
225
200

828 N/mm
4.5 MPa615 N/mm
2.5 MPa

0.5 MPa

300

400

500
600
700
In-plane force [N/mm]

800

900

900

y = 100.84x + 370.98
R² = 0.999

In-plane force

800

In-plane force [N/mm]

400

Linear (In-plane force)

700
600

Additional actionof prestress

500
400
300

Built-in restraint giving rise tocompressive membrane action

200
100
0

0

1

2

3

TPL [MPa]

4

5

(a)
(b)
Fig. 5. CMA: (a) Relationship between the punching load and the in-plane force for varioustransverse prestressing levels; (b) In-plane forces due to the initial prestressing and due to thebuilt-in restraint (compressive membrane force, CMF) in the model bridge deck slab.
5. References
Amir. 2014. Compressive Membrane Action in Prestressed Concrete Deck Slabs. PhD thesis.
Delft University of Technology, Delft, the Netherlands, 282 pp.
Amir, S., van der Veen, C., de Boer, A., and Walraven, J.C. 2016. Experiments on Punching
Shear Behavior of Prestressed Concrete Bridge Decks, ACI Structural Journal, 113 (03), 627-636.
Amir, S., van der Veen, C., de Boer, A., and Walraven, J.C. 2019. Numerical investigation of thepunching shear capacity of transversely prestressed concrete deck slabs, fib Structural Concrete,
DOI: 10.1002/suco.201800285.
DIANA. FX+. 2012. User’s Manual 9.4.4. Delft: TNO Building and Construction Research.

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Modelling of Bend Corners of FRP Reinforcement for Concrete Structures
Considering Material Anisotropy
Yuanzhang Yang1 and Weijian Zhao2*
1: Department of Architectural and Civil Engineering, Zhejiang University, Hangzhou, China;email:yy701@zju.edu.cn
2: Department of Architectural and Civil Engineering, Zhejiang University, Hangzhou, China;email:zhaoweijian@zju.edu.cn*: corresponding author

Keywords: FRP reinforcement; FE modeling; bent corners; material anisotropy
Abstract: The bend corner of Fibre-reinforced polymer (FRP) reinforcement has been shown significantly
lower tensile strength than the straight portion. However, the premature failing mechanism at bent corners
has not been fully understood. This paper presents a numerical investigation into the bent-corner strengthof FRP stirrups with rectangular cross-sections based on a refined Finite Element modelling approach
using Hashin’s failure criterion. The models are established and validated against existing experimental
research. Analyses of the mechanism of strength reduction at the bent corners of FRP stirrups withrectangular cross sections are conducted. It shows that the tensile stress developed at the bent corners are
extremely higher than the average tensile stress in the elastic range, which could be caused by the material
anisotropy, concrete compression and bonding. Their contributions to the stress concentration at the bent
corners vary along with the increasing average tensile stress.
1. Introduction
Fiber-Reinforced-Polymer (FRP) has been used in the construction industry for concrete structures (RC)for over four decades. Specifically, to resist shear force, FRP stirrups are used for concrete structures.
However, the high strength of FRP materials cannot be fully utilized at the bent corners of the FRP
stirrups. Numerous experimental investigations have been conducted to study the bent corner strength of
FRP stirrups. It is found that the bent corner strength is only 24%-85% of the uniaxial tensile strength [1,2]. Based on the experimental research, several empirical models of the bent-corner strength were
proposed for the prediction of the reduced bent-corner strength [1-7]. However, to date, due to theempirical nature of the prediction models, the models’ validities vary depending on the size of the databaseused for calibration. The failure mechanism of the bent corners of FRP reinforcement has not been fully
understood. This paper established refined FE models simulating the bent corner of FRP reinforcementwith rectangular cross-sections based on Hashin’s failure criterion [8]. The models were verified against
an experimental research in the literature and the mechanism of the reduced strength at bent corners wasanalyzed focusing on the stress concentration and damage accumulation.
2. Reference Experimental Research and Finite Element Modelling
The reference tests were adopted from the experimental work of Spadea, Orr [7], where six groups of 18
specimens were tested by pulling the concrete blocks by the testing machine with the steel bolts until
failure, as shown in Fig. 1. Six refined FE models were established for the corresponding groups of test
specimens based on the ABAQUS platform, as shown in Fig. 2. To reduce the calculation expenditure,quarter models were adopted for calculation, capitalizing on the symmetry of the specimens in the
direction of the applied load (Z axis) and the width of FRP stirrups (Y axis).
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Fig. 1. Test setup of FRP reinforcement specimens (a) and design configurations (b) [7]Fig. 2. Discretization and boundary condition of the simulation models
For concrete material modelling, the constitutive relations from the Chinese design code for concrete
structures GB50010 [9] was adopted and the material properties of concrete are shown in Table 1. A bilinear progressive damage model based on the Hashin [8] failure criterion was adopted to model the FRP
stirrups. The material properties of the FRP stirrups are shown in Table 2.
Table 1. Concrete material properties
Concrete
Value

(MPa)

38

2.9

(MPa)

33

(GPa)

Table 2. Calculated material properties of FRP stirrups
Specimens

(MPa)

(MPa)

(MPa)

(MPa)

(MPa)

(MPa)

(GPa)

(GPa)

(GPa)

(GPa)

(GPa)

Reference

1378

950

40

125

97

45

129.0

5.6

0.32

1.33

1.33

1.86

1(t=0.7mm)
2(t=1.0mm)
3(t=1.3mm)

1537
1503
1484

1060
1036
1023

45
44
43

139
136
135

108
106
104

50
49
48

109.5
108.7
107.7

6.6
6.6
6.7

0.32
0.32
0.32

1.33
1.33
1.33

1.33
1.33
1.33

1.86
1.86
1.86

4(t=1.6mm)
5(t=1.9mm)

1487
1426

1025
983

43
41

135
129

105
100

49
47

107.1
106.6

6.7
6.8

0.32
0.32

1.33
1.33

1.33
1.33

1.86
1.86

6(t=2.2mm)

1384

954

40

126

97

45

105.8

6.8

0.32

1.33

1.33

1.86

3. Model Validation
3.1. Ultimate failure load
The validity of the FE modelling was verified by the comparisons of ultimate capacity . The simulated
and tested failure loads are compared in Table 3. The comparisons show that FE models established in thiswork can effectively simulate the failure load of FRP reinforcement with rectangular cross-sections at bent
corners. For all six sets of specimens, the discrepancy of
between the simulations and tests varied
from 1.10% to 7.40%.
Specimens

Table 3. Failure load comparison
1

2

3

4

5

6

Tested Pfb (kN)

4.1

6.38

9.19

11.91

14.06

16.01

Simulated Pfb (kN)

4.22

6.45

9.87

11.75

13.3

15.16

Error (%)

2.92

1.1

7.4

1.34

5.41

5.31

Mean error (%)

3.9

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4. Failure Mechanism of the FRP Bent Corners
4.1. Tensile stress distribution and damage accumulation
Along with the increasing applied load, Fiber Tension (FT) damage was observed on the first element at
the inner perimeter of the bent corner, adjacent to the straight part. It was observed that the initially FTdamaged elements carried higher tensile stress than the remaining outer elements along the thickness
direction, as shown in Fig. 3 (left).

Fig. 3. Tensile stress distribution of the bent corners before (left) and after (right) FT damage appearanceWith increasing load, the FT damage maintained at the same locations until FT damage factor reached 1.0,
indicating the fibers were fully damaged under tension, as shown in Fig. 3 (right). The location of tensilestress concentration shifted from the initially FT damaged element to outer layers of elements.
Consequently, the outer layer elements began to endure FT damage. The FT damage accumulated alongwith the increasing higher applied load. When the specimens reached their ultimate capacities, all fivelayers of FRP elements along the thickness direction endured FT damage, with approximately half ofwhich were fully damaged. The location of the tensile stress concentration was close to the outer perimeterof the FRP bent corner.
4.1. Contributing factors to the tensile stress concentration
The tensile stress concentration of the FRP stirrup specimens was the critical reason for the FT damage
accumulation and resulting failure of the bent corners. This stress concentration could be caused by two
reasons: 1) bending moment at bent corner and 2) the anisotropy of the FRP material. To quantify the
proportions of stress concentration contributed by FRP anisotropy and the applied bending moment at thebent corners in the elastic range, additional FE models were established. Model I had the same geometricaldimensions and friction coefficient (μ = 0.2) as the model of Group 3 (Error! Reference source not found.)but the material properties of the reinforcement were set as isotropic- elastic, which had the same tensile
strength and elastic modulus of FRP in the axial direction. Model II had the same setting as Model I,
except for being set as zero to exclude the influence of bonding. The maximum tensile stresses for Group 3,
Model I and Model II were compared at 100MPa, 200MPa and 300 MPa of average tensile stress, as
shown in Figure 4. The difference between the maximum stress and average stress for Group 3 could be
calculated as the summation of the contributions from material anisotropy and the applied bending
moment (including compression from concrete and friction). For Model II, the difference between Smax andSave could be considered as the contribution of concrete compression. Whilst the difference of Smax between
Model I and Model II and that between Group 3 and Model I could be considered as the contributions offriction and material anisotropy, respectively. In the elastic range, the material anisotropy accounted for
45% to 55% of the tensile stress concentration, followed by compression from concrete accounting for
34% to 48% and friction accounting for 7% to 13%.
Table 4. Contributions to tensile stress concentration in elastic range
(MPa)
100
200
300

Group 3
(MPa)
262
465
691

(MPa)
Model I
(MPa)
172
338
517

Model II
(MPa)
155
303
490

Difference of
and
(MPa)
162
265
391

315

Material
anisotropy (%)
55%
48%
45%

Contributions
Compression of
concrete (%)
34%
39%
48%

Bonding
(%)
11%
13%
7%

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Fig. 4. Tensile stress distribution of model of Group 3, Model I and Model II
5. Conclusions
This paper presents a numerical investigation into the failure mechanism of the FRP bent corners
with rectangular cross sections. The following conclusions are supported by this work:
1. The modelling approach developed in this work can effectively simulate the failure of FRP
bent corners, verified against tested data.
2. The tensile stress concentration at the bent corners of FRP reinforcement was the major reason
for the initial damage occurrence. The resulting damage developed from the inner perimeter ofthe bent corner to the outer perimeter with increasing applied load.
3. The tensile stress concentration was impacted by the compression from concrete, bonding andthe material anisotropy of FRP, the proportions of which vary along with the average tensilestress
6. Acknowledgement
The authors acknowledge and are grateful for the support of the National Natural Science
Foundation of China (52208216) and the Zhejiang Provincial Natural Science Foundation(LQ21E080021), who funded and resulted in this study.
7. Reference
1. Ahmed, E.A., et al., Bend Strength of FRP Stirrups: Comparison and Evaluation of Testing
Methods. Journal of Composites for Construction, 2010. 14(1): p. 3-10.
3. El-Sayed, A.K., E. El-Salakawy, and B. Benmokrane, Mechanical and structuralcharacterization of new carbon FRP stirrups for concrete members. Journal of Composites forConstruction, 2007. 11(4): p. 352- 362.
4. Imjai, T., et al., Strength Degradation in Curved Fiber-reinforced Polymer (FRP) Bars Used asConcrete Reinforcements. Polymers, 2020. 12(8): p. 1653.
5. Lee, C., M. Ko, and Y. Lee, Bend strength of complete closed-type carbon fiber-reinforced
polymer stirrups with rectangular section. Journal of Composites for Construction, 2014. 18(1): p.
04013022.
6. Nakamura, H. and T. Higai, Evaluation of Shear Strength on Concrete Beams Reinforced withFRP.
7. Concrete Library JSCE, 1995. 1995(26): p. 111-123.
8. Nagasaka, T., H. Fukuyama, and M. Tanigaki, Shear performance of concrete beams
reinforced with FRP stirrups. ACI Special Publication, 1993.

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9. Spadea, S., J. Orr, and K. Ivanova, Bend-strength of novel filament wound shear
reinforcement. Composite Structures, 2017. 176: p. 244-253.
10. Hashin, Z., Failure Criteria for Unidirectional Fiber Composites. Journal of Applied
Mechanics, 1980, 47(2): p. 329-334.
12. CABR, Code for design of concrete structures, in GB50010-2010. 2016, Chinese Academyof Building Research: Beijing.

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Water Pressure Variation on Porous Calcium-Silicate-Hydrates
Using Molecular Dynamics
Shota Takinami 1*, Ryo Yoshida2, and Ryo Kobayashi31: Nagoya Institute of Technology, Aichi, Japan; email: clz16025@ict.nitech.ac.jp
Nagoya Institute of Technology, Aichi, Japan; email: yoshida.ryo@nitech.ac.jp
3: Nagoya Institute of Technology, Aichi, Japan; email: kobayashi.ryo@nitech.ac.jp
2:*: corresponding author

Keywords: Molecular dynamics; calcium silicate hydrate (C-S-H); simulation
Abstract:
Many experiments have shown that water in the pores of concrete plays an important role in thecapabilities and deterioration process of cement paste, but there are many unclear issues about thecalcium silicate hydrate (C-S-H) of its microstructure. One of these concerns is the change involume of C-S-H with water content. Ca/Si ratio of the crystal structure has a significant effect on
the volume change of C-S-H, which varies with the type of concrete. Ordinary Portland Cement(OPC) has a large Ca/Si ratio but a small volume change. There are many experiments and analysis
that ignore the volume change due to drying and wetting. Recently, however, environmentallyfriendly Blast Furnace Slag (BFS) has increasingly been in use. It has a smaller Ca/Si ratio thanOPC, and its volume change is larger. It is therefore necessary to reconsider the assumptions ofmany experiments. In this study, the molecular dynamics (MD) method was used to understandthe physical properties of C-S-H. Water molecules were placed between two C-S-H layers to checkwater migration. The pressure in the C-S-H layer was investigated by varying the number densityof water molecules, interlayer distance, and temperature, and it was confirmed that when thenumber of water molecules was constant, a water column was formed by the intermolecular forcesof water molecules. This water column exerts pressure in the cohesive direction on the C-S-H layer.
As the number of water molecules is further increased, the pressure in the direction of cohesion istransferred to the direction of expansion to the extent that the space between layers is filled. It issuggested that this pressure-induced volume change process is a factor that should be consideredin many studies and experiments.
1. Introduction
Portland cement is the second most widely used industrial product on earth after water, and one of
the most important industrial products supporting the society. Although microcrystalline C-S-H,
the major hydrate of Portland cement, is the main hydrate that determines the strength and Young's
modulus of concrete, pore structure, and mass transfer resistance that govern the durability ofreinforced concrete, the composition, structure, and other details of C-S-H have not been clarified.
The calcium/silica ratio (Ca/Si) in the crystalline structure of C-S-H has a significant influence on
the volume change, which depends on the type of cement. Ordinary Portland cement (OPC) has alarge Ca/Si ratio but a small volume change. As a result, there are many experiments and analysis

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that ignore the volume change caused by drying and wetting. However, in recent years, the use ofenvironmentally friendly blast furnace slag (BFS) has widely increased, which has a smaller Ca/Si
than OPC, and thus the volume change due to water content is greatly affected.
2. Method
2.1 C-S-H model
In this study, the Tinf_14sc model is used for a multimeric structure model of Tobermorite
minerals proposed by Richardson, and reproduce C-S-H by sandwiching water molecules between
them from above and below. The Tinf_14sc model is a monoclinic system classified in the spacegroup 'P21/c' and consists of angles α = 90.000°, β = 95.500°, and γ = 90.000°. The box is 11.300
Å × 7.300 Å × 28.800 Å. However, in order to make the simulation easier to handle, the box 11.300
Å × 7.300 Å × 28.667 Å, was made consisting of angles α = 90.000 °, β = 90.000 °, and γ =90.000 °. The SPC model, a simple point charge model, was used for the water molecules.
Table 1. Various parameters of the SPC modeli
O（g/mol）
15.994H（g/mol）
1.008

𝑞𝑞𝑂𝑂 [C]-0.82

𝑞𝑞𝐻𝐻 [C]
0.41

r（Å）
1

θ（°）
109.5

Table 2. SPC Model Lennard Jones
Potential Parameters
j

Ow
Hw
Hw

Ow
Ow
Hw

𝜎𝜎𝑖𝑖𝑖𝑖 （Å）
3.16412-ℇ𝑖𝑖𝑖𝑖 （Kcal/mole）
0.161776-

2.2 Molecular dynamics (MD) Method
Molecular dynamics (MD) is a method that reproduces molecular behavior derived from Newton's
second law, in which the particles constituting the system are modeled as the simplest mass points,
and the time evolution of the position and velocity of each atom is shown in Equation (1).
𝑚𝑚𝑖𝑖

𝜕𝜕 2 𝑟𝑟𝑖𝑖
𝜕𝜕𝑈𝑈=−
𝜕𝜕𝑡𝑡 2
𝜕𝜕𝑟𝑟𝑖𝑖(1)where 𝑚𝑚𝑖𝑖 and 𝑟𝑟𝑖𝑖 are the mass and position vector of atom i, respectively, and U is a function of
the potential energy of the system. In this study, equation (2) was used for U.
12

𝜎𝜎𝑖𝑖𝑖𝑖𝑈𝑈 = � � 4ℇ𝑖𝑖𝑖𝑖 �� �
𝑟𝑟𝑖𝑖𝑖𝑖
𝑖𝑖

𝑗𝑗>𝑖𝑖

6

𝜎𝜎𝑖𝑖𝑖𝑖𝑞𝑞𝑖𝑖 𝑞𝑞𝑗𝑗− � � � + � � 𝑘𝑘
𝑟𝑟𝑖𝑖𝑖𝑖𝑟𝑟𝑖𝑖𝑖𝑖
𝑖𝑖

𝑗𝑗>𝑖𝑖

(2)

where 𝑟𝑟𝑖𝑖𝑖𝑖 is the distance between two atoms, k is the proportionality constant that defines
Coulomb's potential, and 𝑞𝑞𝑖𝑖 is the atomic charge. The first term is the Lennard-Jones potential
describing the repulsion due to the overlap of atoms and the attraction due to the van der Waals
interaction. The second term is the Coulomb potential, which describes the electrostatic interaction.
The following Adams et al. equation is used to determine the charge�⃓𝑉𝑉 �𝑋𝑋 �𝑁𝑁⃓∑𝑗𝑗 𝑖𝑖𝑖𝑖 𝑗𝑗 𝑋𝑋𝑋𝑋⃓
𝑉𝑉𝑖𝑖𝑖𝑖 (𝐴𝐴𝑖𝑖 ) ⃓�𝑛𝑛𝑋𝑋𝑋𝑋⃓
𝑞𝑞𝐴𝐴𝐴𝐴 =,⃓�𝑛𝑛𝐴𝐴𝐴𝐴 ⃓⃓ ∑𝑖𝑖 𝑉𝑉𝑖𝑖𝑖𝑖 (𝐴𝐴𝑖𝑖 )𝑁𝑁𝐴𝐴𝑖𝑖
�𝑛𝑛𝐴𝐴𝐴𝐴⎷�
𝑉𝑉 (𝐴𝐴 )𝑁𝑁⃓∑𝑖𝑖 𝑖𝑖𝑖𝑖 𝑖𝑖 𝐴𝐴𝐴𝐴⃓⃓
𝑉𝑉𝑖𝑖𝑖𝑖 �𝑋𝑋𝑗𝑗 � ⃓�𝑛𝑛𝐴𝐴𝐴𝐴⃓
𝑞𝑞𝑋𝑋𝑋𝑋 =⃓ 𝑉𝑉 �𝑋𝑋 �𝑁𝑁
�𝑛𝑛𝑋𝑋𝑋𝑋 ⃓⃓∑𝑗𝑗 𝑖𝑖𝑖𝑖 𝑗𝑗 𝑋𝑋𝑋𝑋�𝑛𝑛𝑋𝑋𝑋𝑋⎷
319(3)

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where 𝐴𝐴𝑖𝑖 is the i-th cation, 𝑋𝑋𝑗𝑗 the j-th anion, 𝑉𝑉𝑖𝑖𝑖𝑖 the ideal charge of the element, N the number ofatoms in the system, and n the principal quantum number. Instead of the Lennard-Jones potential,
P.A. Bonnaud et al. adopted the mathematical form used in the PN-TArZ model shown inequations (4) and (5).
5𝑈𝑈𝑖𝑖𝑖𝑖 �𝑟𝑟𝑖𝑖𝑖𝑖 � = 𝐴𝐴𝑖𝑖𝑖𝑖 exp�−𝑏𝑏𝑖𝑖𝑖𝑖 𝑟𝑟𝑖𝑖𝑖𝑖 � − � 𝑓𝑓2𝑛𝑛 �𝑟𝑟𝑖𝑖𝑖𝑖 �2𝑛𝑛

𝑓𝑓2𝑛𝑛 �𝑟𝑟𝑖𝑖𝑖𝑖 � = 1 − �

𝑚𝑚=0

𝑛𝑛=3
𝑚𝑚

𝑖𝑖𝑖𝑖

𝐶𝐶2𝑛𝑛
𝑟𝑟𝑖𝑖𝑖𝑖2(4)

�𝑏𝑏𝑖𝑖𝑖𝑖 𝑟𝑟𝑖𝑖𝑖𝑖 �exp�−𝑏𝑏𝑖𝑖𝑖𝑖 𝑟𝑟𝑖𝑖𝑖𝑖 �
𝑚𝑚!

(5)

This was used to represent the intermolecular forces in the repulsive and attractive portions of theinteraction between the Tobermorite system C-S-H. Since the software used in this study,
LAMMPS (https://www.lammps.org/), cannot use the PN-TArZ model as it is, the PN-TArZmodel was approximated to the Lennard-Jones potential.
Table 3. Charge of each atom in the Tinf_14sc
model

Table 4. Lennard Jones for Tobermorite system
Potential Parameters

T∞_14sc

Ca

Si

O

i

j

𝑁𝑁

8

12

36

𝑉𝑉𝑖𝑖𝑖𝑖

2

4

2

𝑛𝑛

4

3

2

𝑞𝑞

1.19392.7573-1.18425

Ca
Si
H
Ca
Si
O

Ow
Ow
Ow
Hw
Hw
Hw

𝜎𝜎𝑖𝑖𝑖𝑖 （Å）

3.211577733
3.844182653
2.66
2.65
3.3979572
3.27333ℇ𝑖𝑖𝑖𝑖 （Kcal/mole）
0.0993886
0.01794068
0.1272912
0.0675658
0.00959296
0.163726

3. Results
3.1 Column of water
The number of water molecules was increased, and when the density exceeded a certain number,a column of water was formed so that water molecules adsorbed on the upper and lower walls werebound together.

(a)
(b)
(c)
(d)
(e)
Fig. 1. Formation of water column: number density (/𝑚𝑚3 ) (a) 0.0121, (b) 0.0181, (c) 0.0217, (d)0.0301,(e) 0.0362
Time averages of at least 100 times were performed. Standard errors were calculated to accountfor variations in the estimated quantities. Number density of water molecules and the pressure in
the direction of expansion are shown when H=10 Å and T=300 K. Error bars indicate standard
errors.

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When a column of water is formed, pressure in the direction of cohesion tends to occur. When thenumber density of water molecules fills the interlayer, the pressure shifts from the direction ofcohesion to the direction of expansion.

Fig. 2. Number density of water moleculesand pressure on the C-S-H wall inthe direction of expansion

Fig. 3. Temperature change, number densityand pressure

3.2 Relationship between temperature and pressure
Under the temperature T=330K, cohesive pressure due to the column of water was observed. The
lower the temperature, the greater the force in the direction of cohesion. This is because the lower
the temperature, the lower the thermal energy of water molecules, the less the water column is
disrupted, and the greater the attraction between water molecules.
In the real external environment, water at temperatures of 270K is solid, but in this simulated
environment, water molecules are fluid and not individuals even at temperatures of 270K. In the
microscopic world, temperatures such as melting points are often different from the actual external
environment
4. Conclusion
The following was clarified(1) As the number density of water molecules between C-S-H layers is increased, water columns
are formed above a certain interlayer distance, which exert inward cohesive pressure on C-S-H
walls.
(2) The lower the temperature, the greater the pressure in the cohesive direction due to the watercolumn.
(3) If the number density of water molecules is made larger than the number density at which watercolumns are formed, the interlayer is eventually filled with water molecules. The pressure in the
direction of cohesion is then transferred as the force in the direction of pressure.
5. References
P.A Bonnaud, Q.Ji, B Coasne ,R.J-M.Pellenq, and K.J.Van Vliet : Thermodynamics of Water
Confined in Porous Calcium-Silicate-Hydrates, Langmuir 2012, volume 28, issue 31, pages
11422–11432, June.2012

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Stefan Adams, R.Prasada Rao:High power lithium ion battery materials by computational design,
physica status solidi applications and materials science, volume 208，issue 8,pages 1746-1753,
August.2011
Ian G. Richardson: Model structures for C(A)-S-H(I), Acta Crystallographica Section B,
B70，pp903-923, Dec.2014322

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Real Imperfections of Longitudinally Stiffened Plates Compared to the
Proposed Imperfections in EN 1993-1-5
N. Thomas1*, A. Weinhuber1*, M. Mensinger1, C. Holst1, and J. Ndogmo1
1: Technical University of Munich, Munich, Germany; email: nadine.thomas@tum.de,a.weinhuber@tum.de, mensinger@tum.de, christoph.holst@tum.de, ndogmo@mytum.de*: corresponding author

Keywords: steel bridges; imperfections; laser Scanning; buckling behavior; stiffened plates
Abstract: For the buckling verification of longitudinally stiffened plates according to EN 19931-5 [1] using the Finite-Element Method, imperfections can be considered either asgeometrically equivalent imperfections or with 80% of the geometric imperfections according to
DIN EN 1090-2 [2] and additionally structural imperfections according to Annex C [1]. The
resulting values are often contradictory. Therefore, to verify this regulation, imperfections of real
bridge components were measured. The measurements were carried out at the Technical
University of Munich during the replacement construction of the Thulba viaduct. The measured
imperfections are compared with the different imperfection regulations for buckling verificationaccording to EN 1993-1-5 and the regulations for manufacturing tolerances according to DIN EN
1090-2. The investigations aim to clarify whether the assumptions of the standards are
sufficiently defined and whether the regulations are consistent with the execution of real bridgecomponents.
1. Introduction
The replacement of the Thulba viaduct is a steel-composite bridge with a box section in northern
Bavaria, which was built in 2021 using the incremental launching method. The bridge has a totalof 7 spans of 45 m – 55 m – 70 m – 90 m – 75 m – 70 m – 55 m and a final length of 460 m.
Large-scale buckling tests were carried out at the Technical University of Munich in 2017 inframe of this new bridge construction. Questions concerning the design of the longitudinal
stiffened panels under biaxial loading during launching, such as the failure of the single panel
under lateral load, the deflection of the longitudinal stiffeners, the load introduction andrecommendations for maximum possible manufacturing deviations were clarified here. The
results were published in [3] and [4]. In addition, the geometry of several panels of the newbridge were additionally measured using 3D laser scanning during construction. The
measurement takes place at a fixed defined bridge section on both sides between four transverse
bulkheads. Laser scans are performed in this bridge section at three different displacement times.
In addition, the temperature influence on the component was investigated by further scans (dayand night measurement). This paper presents the first measurement results and compares themwith the imperfections from current regulations for the buckling verification and the
manufacturing tolerances. Furthermore, only the panels between the longitudinal stiffeners areanalyzed in this paper. This allows a basic global comparison and a comparison with theimperfection regulations for sub- or single panels.

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2 Existing Regulations for Imperfections
Especially for buckling verification, the correct assumption of imperfections is essential for thedesign of the components. On one hand, the imperfection must be chosen sufficiently large toconsider imperfections from the production process conservatively enough, on the other hand, it
must not deviate too much from the real construction in order to guarantee an economic
construction. In addition it has to cover structural imperfections from residual stresses. For
buckling verification, the imperfection which has to be considered in a FE calculation isregulated in EN 1993-1-5 Annex C. Imperfections used in FE calculations should normally
include both geometric and structural imperfections (residual stresses). In cases where moreaccurate separate approaches of geometric and structural imperfections are not possible,
geometric equivalent imperfections can be used. Geometrical imperfections may be applied in
the form of the modal buckling figure. A value corresponding to 80% of the geometricmanufacturing tolerances is recommended. The manufacturing tolerances are specified in
DIN EN 1090-2. Structural imperfections may be applied with a residual stress distribution
whose course and amplitude can be expected on average from the manufacturing process. As arule, the direction of the imperfections is to be set in such a way that the lowest bearing capacity
is achieved. For geometrical equivalent imperfections, the approaches in Table C.2 and Fig. C.1
of EN1993-1-5 may be used. When considering the interaction of different imperfections, aleading imperfection is usually to be selected, this is applied at 100% to the modal buckling
figure, the accompanying imperfections may be reduced to 70%. If the buckling verification is
performed using the reduced stress method and the column-like behavior occurs, theimperfection is considered in form of an imperfection coefficient α when determining the
reduction factor χc. The imperfection coefficient α shall normally correspond to the buckling
curve a with α = 0.21 for unstiffened plate sections. For stiffened plate sections, α is usuallyreplaced by the increased value αe to account for larger imperfections of welded plates. To
compare the measurements with the manufacturing tolerances given in [2], No. 7 of Table B.1
from [2] is used as a substitute, since no separate regulations are given for deviations for the
stiffened plate and the single panels. This regulates the deviation via the web height b andapplies to the web curvature of welded sections.

min (a/200; b/200)
Imperfections for sub or single panels according toEN 1993-1-5 Annex C

i= ± b/200 für b/t ≤ 80
Manufacturing tolerances according to
DIN EN 1090-2 No. 7 Tab.B.1

Fig. 1. regulations on imperfections

3. Metrological Support by 3D Laser Scanning
3D laser scanners capture the geometry of the environment in a high spatial resolution using alaser beam that is deflected in any direction by a rotating mirror. The result is a 3D point cloud
which allows the panels to be analyzed for deformations. In order to be able to initially detect the
manufacturing imperfections of the construction component, various factors must be taken into

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account in the metrological support. On one hand, it is important to ensure absolute point
accuracy, and on the other hand, the webs are measured with a high resolution of max. 5 mm
point spacing. The conditions are realized by measuring with the HDS7000 laser scanner from
Leica with a setting of the resolution of the scanner of 3.1 mm to 10 m. The single point
accuracy of the scanner can be assumed to be <1 mm. For complete coverage of the webs, it isnecessary to select multiple viewpoints for the scanner to a) minimize shadowing by thestiffeners, b) ensure point accuracy (avoid glancing intersections), and c) ensure sufficient
resolution. The data acquired from different viewpoints are pre-registered via sprayed targets andtransferred to uniform coordinate system as best as possible using the Iterative Closest Point
Algorithm (ICP Algorithm). A registration uncertainty of the individual scans <1 mm can be
determined. The first scans are made when the bridge is still in the assembly yard. The firstimperfections of the component can already be derived here in the load-free state and allows adeformation analysis to manufacturing processes. These can be calculated from the registeredpoint clouds. The imperfections are to be considered as the perpendicular distance of theindividual points to an ideal plane. Since the material thickness can also vary within a web due tothe constructional design, the ideal planes must be estimated locally from the point cloud.
Otherwise, a change in material thickness at the joints would show up as apparent deformation.
In addition, the calculated imperfections of all individual points are interpolated to a uniform
5 mm grid to minimize any measurement noise that may be occurring. In order to make theimperfections accessible to all project participants, a special Matlab tool is provided in which thedetermined imperfections can be visualized and retrieved for individual points (see fig. 2). Figure
2 shows an example of the determined imperfections of the field 40-39 (west side) in the loadfree condition, which already show amounts of up to -6 mm.
4. Comparison of Measurement Results with Different Standards
When comparing the measured deviations perpendicular to the plate with data from [1] and [2],maxima at the edges of the plate are not taken into account, since welds and transitions to thenext component are also located here, which may be included in the results. In addition, forglobal buckling, the maximum deviation of the inner area of the entire panel is decisive; for local
buckling, the deviation of the inner areas of each sub panels are decisive. Figure 2 shows one 3D
scan of a plate as an example with the respective names of the dimensions.

Fig. 2. Example of evaluation of one panel with designations
The maximum measured deviation of each panel is compared with the data of the global
geometric equivalent imperfection according to EN 1993-1-5:2019 Annex C [1] and withreference to Table B.1 No. 7 of DIN EN 1090-2 [2] to the entire web height b. The measured

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valued are significantly smaller for every panel (see table 1). The external dimensions of the web
and the deviations from the 3D scan, as well as the values according to [1] and [2] aresummarized in Table 1. Table 2 summarizes the height of each sub panel of each web. For eachweb, the sub panel with the maximum deviation from the 3D scan is given and then comparedwith the data of the local geometric equivalent imperfection according to EN 1993-1-5:2019
Annex C [1] and with Table B1 No. of DIN EN 1090-2 [2] with reference to b to the respective
sub panel. In this comparison, the values for each sub panel are slightly exceeded.
Table 1. External web dimensions an global maximum deviationWeb
40-39 West

Length a[mm]3977

40-39 Ost

3968

41-40 West

3977
3968

41-40 Ost
42-41 West42-41 Ost

Height b Max. deviation in the sub[mm]panel [mm]
3445
2,90
3169
2,50

Min ( a/400;
b/400) [1]8,61± b/200 [2]
17,23

7,92

15,85

-4,00

8,60

17,20

3163

2,70

7,91

15,82

3394
3118

-3,10

8,49

16,97

3,10

7,80

15,59

3439

3977
3968

Table 2. Web dimensions and local maximum deviationWeb
40-39 West40-39 Ost
41-40 West41-40 Ost
42-41 West42-41 Ost

b1
[mm]
440

b2
[mm]
425

b3
[mm]
600

b4
[mm]
840

Position of the maximum
deviation in the sub-panelb1Max. deviation in the subpanel [mm]2,90Min (a/200;
b/200) [1,2]
2,20

440

425

600

564

b2

2,50

2,13

440

425

600

834

3,00

425

600

558

b3
b2

-4,00

440

2,70

2,13

440

425

600

789

b2

-3,10

2,13

440

425

600

513

b2

3,10

2,13

5. Summary
The 3D Laser scans of the Thulba viaduct, provide important results for the design and executionof steel and steel composite bridges. In this paper it could be shown with a global and local
comparison that the measured values are in a good range. The local comparison with Table B.1
No.7 is not quite correct. In the real component, the stiffeners also deform. The assumption oftable B.1 No. 7 [2] as well as the assumption of local imperfection according to [1] assume a"0 position" of these stiffeners. In the measurement, the deformations of the longitudinal
stiffeners are also taken into account in the sub panels.
6. References[1] DIN EN 1993-1-5:2019-10: Eurocode 3: Design of steel structures –Part 1-5: Plated structural elements;
German version EN 1993-1-5:2006 + AC:2009 + A1:2017 + A2:2019 (October 2019).
[2] DIN EN 1090-2:2018-09: Ausführung von Stahltragwerken und Aluminiumtragwerken –Teil 2: Technische
Regeln für die Ausführung von Stahltragwerken; Deutsche Fassung EN 1090-2:2018[3] M. Mensinger, J. Ndogmo N. Thomas (née Maier), U. Kuhlmann und V. Pourostad: Untersuchungsbericht Nr.
118001: BAB A7, Würzburg - Fulda, BW613a, Talbrücke Thulba, Beuluntersuchungen; Technische
Universität München, MPA Bau Abteilung Metallbau. 2019.
[4] Kuhlmann U., Mensinger M., Thomas N. (née Maier), Pourostad V., Ndogmo J., Schlussbericht AIF IGF Nr:
20455, Längsausgesteifte Beulfelder unter mehrachsiger Beanspruchung“, Universität Stuttgart undTechnische Universität München, 2021

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Structural Damage Prediction with Multi-Level Neural Networks Using
Modal Group Response
Yunfeng Zou1, Xuandong Lu 2*, Xuhui He3, and Chenzhi Cai4
1

: School of Civil Engineering,
yunfengzou@csu.edu.cn
2: School of Civil Engineering,
xuandonglu@csu.edu.cn
3: School of Civil Engineering,
xuhuihe@csu.edu.cn
4: School of Civil Engineering,
chenzhi.cai@csu.edu.cn

Central South University, Changsha, China; email:Central South University, Changsha, China; email:Central South University, Changsha, China; email:Central South University, Changsha, China; email:*: correspondence author

Keywords: large dimension; closely spaced modes; modal group; multi-level neural networks
Abstract: The system with large dimension and closely spaced modes in civil structures affects
the development of damage prediction. In this paper, the empirical mode decomposition withintermittence criterion is used to extract the modal group including several closely spaced modes,
and the multi-level neural networks are applied to reduce the damage searching dimension.
Before practical application, the modal group strain energy is constructed as features from themodal group response to train the feedforward neural network to locate the damaged region, andthe reconstructed modal group is used as features to train several neural networks for internal
damage prediction of each substructure. Structural damage will be first located it thesubstructural level and then be quantified at the elemental level. The numerical study is
performed to verify the feasibility of the proposed method.
1. Introduction
Damage prediction based on modal analysis is widely performed. However, the difficulties in its
practical applications include two aspects. On one hand, when the response includes modal
responses with closely spaced modes, the accuracy of extracting modal parameters will
significantly decrease, thereby affecting the accuracy of damage prediction. On the other hand,
this method is essentially an inverse problem that requires iterative solutions on structural
matrices. However, large-scale structures are often with large dimensions, which leads to a highconsumption of computational resources and time in matrix operations. As the number ofsearching elements increase, the computational efficiency and accuracy of the inverse problem
significantly decrease. For the former, model reduction and multi-level model neural networks(NNs) could be introduced to reduce matrix computation and damage searching dimensions,
respectively. For the latter, a potential improvement strategy is to construct modal group strainenergy (MGSE) and reconstructed modal group (RMG) from modal group response (MGR) as

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features of NNs, which is used for substructural level and elemental level damage prediction
respectively, so as to avoid directly extracting singular modal information. In this paper, thefeasibility of the proposed method is verified by the numerical case of a transmission tower. The
results indicate its superiority over traditional methods.
2. Modal Group Reconstruction
Several modes that are hard to separate into the single-frequency response due to frequency
proximity is set as a modal group. Assuming that Ng modal groups are sorted, and the g-th modalgroup contains the g1-th to gg-th modes, the modal relation between m1-DOFs and m2-DOFs (m1and m2 can be larger than 1) at the g-th mode group can be expressed as:(1)
Where d m g and d m g denote MGR extracted from measurement by empirical mode
decomposition (EMD) (He et al., 2023); φ m g and φ m g are matrix blocks corresponding to m11

2

1

2

DOFs and m2-DOFs under the g-th modal group; the superscript +l and +r represent left inverse
and right inverse, respectively. Only if the number of measured a-DOFs is larger than thenumber of its modes contained in the modal group can φ+lag be effective.
3. Substructure-Level Damage Locating
Based on finite element, assuming that the original structure is divided into Ns substructures,
according to Craig-Bampton fixed interface modal synthesis method (He et al., 2023), theoriginal structure model can be transformed into a super-element model (SEM) with muchsmaller dimension after modal transformation. To reduce the search dimension, the damage isfirst located at the substructural level. The signal “~” is added to the top of the character toindicate SEM terms. The equivalent damage factor θ as labels of the feedforward neural network(FNN) is preset to obtain a set of training data, and the global stiffness matrix is expressed as:(2)
Where Ku is the undamaged global super-elemental stiffness matrix, and K s represents thecontribution of the s-th super-element to the global stiffness matrix.
As the training features of LNN at substructural level, MGSE is defined as:(3)
Where φ and φ+l are the global and measured modal group. Due to the difficulty of providinggmg

global measurement for input, non-measured components in φg is replaced by corresponding
ones in the undamaged SEM. As for the pre-built LNN, the input is expressed as:(4)

Where the part in φ g φ+lmg corresponding to measured DOFs is calculated by Eq. 1, while others are
extracted from the SEM. The LNN receiving the input will provide the probability of damage ofeach substructure. Thecloser the result is to 1, the more likely the substructure is damaged.
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4. Element-Level Damage Prediction
Only damage within single substructure is considered (the most likely condition in the real-timemonitoring). The FNN for the possibility of damage in each substructure will be constructed inadvance. The s-th substructure is used as an example for illustration. Assume that the stiffness
matrix of the s-th substructure is composed of Nes elements and is formed similar to Eq. 2:
Nes

(5)

K(α) = K u +∑θi K i , (-1 ≤ θi ≤ 0)
i =1

Where K s denotes the elemental contribution of the i-th element in the s-th super-element to theglobal stiffness matrix; θi denotes the i-th elemental damage factor as the label. To train the FNN,features expressed as φ m g φ+lm g are calculated by the SEM, where φ+lm g and φm g are corresponding
to measured DOFs inside or near the s-th substructure. According to the substructure-level
locating results, the corresponding FNN used for element-level prediction is enabled. The input
from measurement in this level is expressed as φ φ+l = d d+r .
1

2

2

m1 g

m2 g

m1g

1

m2 g

5. Numerical Case
A tower numerical case shown in Fig. 1(a) example was studied to demonstrate the feasibility ofthe proposed method. The finite element model of the tower is established in ANSYS, with a
total of 808 nodes and 928 elements, including 4824 DOFs. The bottom support of the tower isprone to be damaged, thus only 36 elements at this location are considered for damage prediction.
Base on that, the model is divided into 5 substructures (S1-S5), in which S1 is transformed into asuper-element retaining 30 master DOFs in ANSYS. The transient excitation is applied to thestructure. The Fourier spectrum of the acceleration and 6 modal groups (MGⅠ~MGⅥ) are shownin Fig. 1(b). At the level of Substructure, 4 substructures in the monitoring region are consideredwith 8 equivalent damages (0, -0.02, -0.04, -0.06, -0.08, -0.10, -0.12, -0.14) respectively, andthen 4096 sets including features and labels are calculated in the SEM. Although cases of single
damage are only considered in the following text, a dataset with a wider distribution of damage
may be used to establish an LNN that better fits the inherent mechanism of damage prediction.
At the elemental level, 9 elements in each substructure are considered with 3 elemental damages(0, -0.15, -0.3), and 19683 training sets are obtained from each substructure. Bayesian adaptive
learning framework is used to optimize the hyperparameter in each iteration of NNs. The
iterative process of each NN is shown in Fig. 1(c), in which 4 NNs of element-level prediction
are with the same iteration curve due to the symmetry of the tower, and the final elemental loss ismuch larger than the substructural loss because both the dataset and outputs of the former arelarger than those of the latter.
As the validation of constructed NNs, 4 damage conditions were considered. Preset
damaged elements are marked in Fig. 1(a). Table 1 lists the damage prediction results. In Table
1, substructure-level locating results is shown by the predicted value θˆ of the preset damage,andNes

the absolute value of maximum errors eumax in undamaged elements. Besides, MSE = ∑(θ i −θˆi )2 is
i =1

used to estimate the global accuracy of elemental results. It can be seen that the proposed method
can accurately predict the local damage.

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5 ×10

-1

Amplitude

4
Fy
Fx

MGⅥ

2

S5
E36

MGⅣ

M Ⅱ

25

30

35
40
Frequency (Hz)

45

50

(b)

S4
E42

Substructural level
Elemental level

105
104
Loss

E31

S2

MGⅢ

MGⅠ

0
20

E18

M Ⅴ

3

1
S1

Fourier spectrum
Modal frequency

E25

S3

103
102

Stochastic force
Preset damaged element

1

10
100

(a)

101

102
Step

103

104

(c)

Fig. 1. Information of tower model and NN training: (a) tower model; (b) Fourier spectrum anddefining modal groups; (c) iterative process of NNs
Table 1. Damage prediction results
PresetDamage Damaged
conditions element damage θi
1
E18
-0.1
2
E25
-0.15
3
E31, E42 -0.1, -0.2
4
E36
-0.25

Substructural leveleumax
θˆ-0.2630.048-0.3290.052-0.4700.031-0.4660.054

Elemental leveleuθˆ-0.1130.005-0.1480.012-0.104, -0.188
0.020-0.241
0.023

6. Conclusion
This paper introduced the theory of the proposed damage prediction method with multi-levelneural networks using modal group response, and presented the results of a numerical case toinvestigate its feasibility. Compared to traditional model updating method, NNs performed indamage prediction can construct a direct relationship from features to labels in advance, thus
more timely monitoring can be provided. RMG and MGSE as features of NNs from modalgroups are used to improve the problem of difficulty in predicting damage in the premise of
closely spaced modes. As for data training in the large-dimension structure, model reduction andmulti-level NNs are applied to decrease the calculation consumption. The proposed method isapplied to conditions about single damage, which accounts for the majority in real-timemonitoring. Multiple damages will be studied in the future study.
7. References
He, J., Guan, X. and Liu, Y. 2012. Structural response reconstruction based on empirical mode
decomposition in time domain. Mechanical Systems and Signal Processing, 28, 348-366.
Zhang, C. and Xu, Y. 2017. Multi-level damage identification with response reconstruction.
Mechanical Systems and Signal Processing, 95, 42-57.
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Use of Steel Orthotropic Box Girders for Light Weight Long Span Bridges Analysis and Design of Pedestrian Bridge in Dubai
Tharun John Joseph1*, Anirudh Desai2, Koyya Satyanarayana3, Bogdan Barbulescu4 and Kannaw
Shrivatsa5
1*: Bridge Engineer, Jacobs Engineering Group, Dubai, UAE; email: Tharun.Joseph@jacobs.com
2: Lead Engineer, Jacobs Engineering Group, Hyderabad, India; email: Anirudh.Desai@
jacobs.com
3: Section Manager, Jacobs Engineering Group, Hyderabad, India; email:Satyanarayana.KoyyaVV@jacobs.com
4: Senior Associate Director, Jacobs Engineering Group, Bristol, UK; email:Bogdan.Barbulescu@jacobs.com
5: Project Manager, Jacobs Engineering Group, Dubai, UAE; email:Kannaw.Shrivatsa@jacobs.com*: corresponding author

Keywords: steel structure, structural dynamics, finite element modelling
Abstract: Orthotropic Steel Decks are efficient structural systems capable to increase thedurability, redundance and efficiency of steel structures and therefore reduce the associated selfweight of the structure. Generally, a typical orthotropic steel deck system (OSD) consists of aflat steel deck plate, stiffened by a series of closely spaced longitudinal ribs with support byorthogonal transverse floor beams. In orthotropic box girders, the OSD is an integral part of thesteel box framing and therefore acts as its flange. The use of stiffened steel plate over onedimensional beams as load carrying elements, results in a higher strength-to-weight ratio ofstructural members. This has led to an increase in the use of orthotropic steel decks, especiallyfor long-span bridges. This paper summarises the analysis and detailed design of the recently
completed pedestrian bridge on King Salman bin Abdul Aziz Road. The bridge connects all
corners of a four-way junction with a tram crossing, with the longest span measuring 76m. The
footbridge is located at one of the prime locations in the city and hence, aesthetics andfunctionality was of prime importance to the Employer. The complex geometry, long designspans and space constraints for the substructure were the major challenges in designing thisfootbridge structure. This infinity shaped four-way bridge is made up of orthotropic steel box
girders with 4.8x2.0m trapezoidal cross-sections, with the main curved spans connected at thecentre to provide torsional stability (Fig 1). The top and bottom flanges of the box are made upof thin plates stiffened with trapezoidal closed ribs longitudinally to increase the flexural
buckling resistance of the compression flanges. Equally spaced internal transverse ring-stiffeners
are provided along box girders to retain the fabricated configuration of the girder when subjected
to torsion and avoid girder distortion that would have resulted in additional longitudinal warping
stresses as well as out-of-plane bending stresses in the individual plates of the box sections. The

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structure was modelled and analysed in Midas CIVIL finite element software. A detailed
dynamic analysis was performed to study the complex dynamic forces induced by pedestrians
and determine all critical vibrational modes in longitudinal, transverse and vertical directions. Inhouse design sheets based on AASHTO LRFD were developed exclusively for the project due tothe complex shape and uncommon cross-sections. While the 3D line beam model was sufficientfor section design and studying the global behaviour and vibrational modes, it did not fully
capture the complex load path at the corners and box girder intersections. Hence, a more realistic
3D Finite Element Model using shell plates was created to substantiate the design assumptions
and understand the complex interaction between the bending, distortion and twisting of the deckat the corners and intersections. The 3D model helped to validate the detailed design, obtain arealistic estimate of the deflections and understand the complex load path and therefore avoiding
critical stress concentrations, optimise member sizes and providing the required camber diagrams
along each web of curved box girders.
1. Introduction
Jacobs was appointed by the Contractor (Besix group) as a Design Consultant to design for
construction a pedestrian bridge required to connect all 4 sides of one of the busiest junctions inJumeirah Beach Residence, Dubai as part of the Design and Build Contract. A non-conventional
pedestrian crossing was indispensable at this location due to the high concentration of pedestrian
traffic and congestion due to tram and vehicular movement from all sides. The design and layoutof the bridge were finalized after considering all the requirements and constraints at the junction.
1.1. Architecture
The bridge has a unique infinity shape, composed of 4 steel decks, linking all 4 corners of thejunction for a smooth pedestrian movement in all directions. The 4 decks form a continuous
infinity shape structure that is simply supported on 4 circular concrete piers with conventional
pot bearings placed in each corner. The longitudinal spans are 76m and 72m and the transverse
spans are 48m and 26m. Elevators and staircases are provided in all 4 corners for ease ofpedestrian access.

(a)

(b)

Fig. 1. (a) Birds eye view of the bridge and (b) Typical cross-section of the girders

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1.2. Steel Orthotropic Box Girder
The curved shape and long design spans required an efficient section with high strength-toweight ratio and hence, orthotropic steel deck system integral with the supporting bridge
superstructure framing as a top flange common to both the transverse floorbeams andlongitudinal box girders was chosen. A typical orthotropic box girder is made up of flat, thinsteel plates stiffened by closely spaced longitudinal and transverse ribs. These ribs result inincreased rigidity and considerable material saving compared to conventional steel structures.
2. Analysis & Design
The footbridge was modelled and analysed in Midas Civil and detail designed using design
sheets developed in-house. The loads and structural steelwork design are based on AASHTOLRFD Bridge Design Specifications, 4th Edition (AASHTO, 2007). Basic section andsubstructure design was done based on the results obtained from the 3-D line beam model. 3-Dline beam model was created to understand the overall behaviour of the structure and extract
internal forces and bending moments for designing the steel cross-sections. However, a detailed
3-dimensional solid model was additionally created and analysed to validate the assumptions
regarding stress flow and member design. The cross-section was designed based on the results
extracted from the line beam model. Standard steelwork details (trapezoidal stiffeners, floor
beams, their intersections and required cut-outs, etc.) were designed based on AASHTO LRFD
Level 1 design but clearly supplemented with 3D line beam and full 3D finite element shell plateanalysis models to capture both global and local behaviours.

(a)

(b)

Fig. 2. (a) 2-D Line model and (b) 3-D Plate models created to analyse the bridge.
2.1. Major Challenges & Issues
Reduced plate thickness and complicated weld details possessed a major risk of fatigue cracking
as observed in many orthotropic box girder bridges around the world, especially at the rib-todeck plate joint and that between the transverse and longitudinal ribs (FHWA, 2012). This was
mainly due to the stresses developed due to the in-plane flexure of the transverse ribs combinedwith the out-of-plane twisting of the longitudinal ribs. In addition, the complex shape of thebridge made the sharp corners prone to high concentrations of stresses. The interaction between
the adjacent spans was complex and it was observed that the loads transferred from the long,
curved spans caused significant twisting of the shorter side spans. The curved and side spansintersected at different angles in each of the four corners, which affected the symmetry of thestructure. Hence, special design considerations had to be given to the load path and the internal
flow of stresses in each of these corners to produce an optimum design.

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All these issues were carefully studied to mitigate all possible construction/operational
challenges. Complete Joint Penetration (CJP) welds were used between all structural members to
improve the fatigue strength of welded joints between and the ribs and deck plate (Dung et. al.,2015). At critical welding locations, base metal materials with improved through-thickness
properties and specific welding details were prescribed to mitigate risk of lamellar tearing.
Staircases to the edges of the transverse spans were designed to be isolated from the steelwork
bridge superstructure to avoid the deck torsion effects being transferred to the staircases. A
detailed 3-dimensional plate model of the entire bridge was created to analyse the stress flow, theconcentration of stresses, and effects of dynamic loading and to study the behaviour of all
structural members in detail as well as determining the precamber diagrams for each box girder
web.
2.2. 3-D Finite Element Model and Design Validation
A detailed 3-dimensional shell plate model helped to understand the overall as well as local flowof stresses along the longitudinal and transverse directions of the bridge deck, especially at thefour corners. It was observed that the bending and torque transferred from the longitudinal
curved spans of the bridge resulted in high concentration of stress at the junctions with thetransverse spans. This model helped the design team to fix the plate and rib dimensions and thespacing at the corners. Furthermore, the 3-D model was able to accurately capture the magnitudeof rotation of all the spans based on which, the required camber design information was
produced.

Fig. 3. High concentration of stresses at the corner
2.3. Dynamic Analysis
Pedestrian Dynamics Analysis was specifically complex and required detailed design using
discrete finite element analysis (3D line model) where the steelwork superstructure, concrete
substructure and piled foundations were modelled. This allowed the determination of all criticalpedestrian induced vibration modes in longitudinal, transverse and vertical directions, based on
the vibration analysis flowchart from SETRA (SETRA, 2006).

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1st Mode

2nd Mode

3rd Mode

Fig. 4. Critical pedestrian induced vibration modes
3. Conclusions
1) Orthotropic box girders can be successfully used for light-weight long-span bridges withcomplex shapes. Decks and webs were typically made up of 16-20mm plates and ribs with 810mm plates.
2) 3-dimesnional plate model helped to accurately capture the intricate details with respect tostress flow, concentration of stresses and displacement/rotation of each structural element. This
helped to arrive at optimum member sizes and required camber.
4. References
AASHTO. 2007. AASHTO LRFD Bridge Design Specifications, American Association of State
Highway and Transportation Officials.
AISC. 1963. Design Manual for Orthotropic Steel Plate Deck Bridges, American Institute of
Steel Construction.
Bethlehem Steel Corporation. 1981. Designer’s Guide to Steel Box Girder Bridges.
Dung, C. V., Sasaki, E., Tajima, K. and Suzuki, T. 2015. International Journal of Steel Structures,
15, 299-310.
FHWA. 2012. Manual for Design, Construction and Maintenance of Orthotropic Steel Deck
Bridges, US Department of Transportation Federal Highway Administration.
NSBA. 2005. Practical Steel Tub Girder Design, National Steel Bridge Alliance.
SETRA. 2006. Technical Guide - Footbridges Assessment of Vibrational Behaviour ofFootbridges under Pedestrian Loading, Service D'études Techniques des Routes et Autoroutes.

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Analytical Study of Fracture Behavior of
Post-Tension Anchorage Zone Using FEM
Kyun-Tae Lee1, Jun-Mo Yang2, and Jin-Kook Kim3*
1: Department of Civil Engineering, Seoul National University of Science and Technology, Seoul,
Republic of Korea; email: nome585@naver.com
2: Department of Civil Engineering, Keimyung University, Daegu, Republic of Korea; email:jm.yang@kmu.ac.kr
3: Department of Civil Engineering, Seoul National University of Science and Technology, Seoul,
Republic of Korea; email: jinkook.kim@seoultech.ac.kr*: corresponding author

Keywords: post-tension; anchorage zone; bearing plate; lateral confinement; construction
quality
Abstract: In the post-tension anchorage zone, even if an anchorage design that has been verified
by passing the performance test is used, the anchorage zone or bearing plate may be destroyed if
the prestressing steel strands are tensioned during construction. In this study, load transfer tests
and parametric analysis were performed for lateral confinement variables, possible construction
errors and product defects to identify the causes of the failure of the anchorage zone and bearing
plate during construction. Through comparison with the experimental results, it was verified thatthe finite element analysis model can effectively simulate construction errors and damage to theanchorage device. A parametric analysis was performed with anchorage device tensile strength,
anchor head effective bearing area ratio, spiral reinforcement start position, spiral reinforcement
spacing, spiral reinforcement outer diameter, anchorage device rib damage, and concrete voids
under the ribs as variables. Based on the analysis results, the factors affecting the fracture of theanchorage device were analyzed, and the strength characteristics of the post-tension anchorage
zone according to design and construction parameters were analyzed.
1. Introduction
With post-tension concrete, tension is transmitted to the concrete through the anchorage as it istensioned, inducing high levels of compressive stress and local bursting stress under the bearing
plate of the anchorage. Because anchorage devices are customized products that require a longdevelopment period and are expensive, most studies focus on the strength of the concrete, thesize of the anchorage, and/or details about reinforcement techniques through experiments andnumerical analyses (Kim et al. 2021; Kwon et al. 2018). However, even with the proper designof the anchorage device and anchorage zone, with the performance verified through load-transfer
experiments and various studies, there have been cases of anchorage zone fractures when steel
strands are placed under tension. In addition, considering that anchorage devices are made of castiron, the product may not be uniform, and partial damage may occur during the construction
process or when receiving a load. Hence, research on these types of possible damage is necessary.

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In this study, to investigate the cause of the failure of the anchorage and bearing plate during
construction, load-transfer experiments were conducted to check for possible construction errors
and product defects, and parameter analysis was performed. The finite element analysis model
was verified through a comparison with the experimental results, and a parametric analysis was
conducted using possible construction errors and product defects as variables. Based on theanalysis results, the factors influencing the failure of the anchorage device were analyzed, as
were the strength characteristics of the post-tension anchorage zone according to the design andconstruction variables. Finally, by synthesizing the results of the study, design directions toprevent the failure of the anchorage zone were presented.
2. Load-transfer Experiment
The experiment was performed according to the procedure of EAD 16004-00-0301 (EOTA
2016). Based on the nominal strength of the steel strand, a load between 0.8 Fpk and 0.12 Fpk was
repeatedly applied more than ten times, and when the stabilization criteria were met, an extremeload was measured by applying such a load. Based on the experimental specimen 15H-1
fabricated to meet the design criteria, the starting position of the spiral reinforcement, thestrength of the spiral reinforcement, the pitch of the spiral reinforcement, and the outer diameterof the spiral reinforcement were tested as variables, as shown in Table 1.
Table 1. Test specimen details
Dt
fy
s
D
Pu,EXP
(mm)
(MPa)
(mm)
(mm)
(kN)
15H-1
50
400
50
330
3,353
15H-2
100
400
50
330
3,254
15H-3
50
500
50
330
3,440
15H-4
50
400
70
330
3,129
15H-5
50
500
60
330
3,315
15H-6
50
500
60
300
3,288
Dt= distance from the top surface to the starting position of the spiral reinforcement
s= pitch of the spiral reinforcement
D= outer diameter of the spiral reinforcement
Pu,EXP= ultimate strength from the experiment
Pu,PTI= ultimate strength calculated using the PTI calculation formula

Pu,PTI
(kN)
3,512
3,512
3,512
3,037
3,387
3,040

The modes of failure of the anchorage are generally defined in three types (Kim et al. 2019):
local zone failures due to inadequate concrete strength or reinforcement levels, spreading failuresdue to longitudinal rupture cracks caused by inadequate reinforcement, and failures at boundaries
between reinforced and unreinforced parts. Here, all specimens were fractured in the bearing
plate of the anchorage device shown in Fig. 1 during the experiment. The reduction of theeffective bearing area following the fracture of the anchorage device followed by the local failureof concrete on the lower surface of the anchorage device is defined as a fourth failure mode.
When comparing the experimental results with the PTI bearing strength, the findings weresimilar on average, and it was determined that the method used to calculate the confinement
force of PTI was effective.

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Fig. 1. Fracture of the bearing plate

Fig. 2. Finite Element model

3. Parametric Analysis
Based on the experimental results, the modeling process here utilized conditions identical tothose used with the specimen shown in Fig. 2, and only a quarter of the specimen was modeled
for efficiency of the analysis. The material model for the nonlinear analysis of each member was
defined to be similar to the experimental results, and the contact conditions and boundary
conditions between each member were identical to those in the experiment. The validity wasreviewed by comparing the load-displacement curves in the experimental results and the analysis
results. A parametric analysis was also conducted to determine the effects of variables notreflected in the design formula as well as any construction errors and/or product defects on the
behavior of the anchorage zone.

(a)

(b)

(c)

Fig. 3. Results of parametric studies: (a) effects of the strength of the anchorage device; (b)effects of the product defects; (c) effects of the cavities

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4. Results and Conclusions
The strength of the anchorage device and effective bearing area ratio are the major factors
influencing the failure of the bearing plate of the anchorage device, and when the bearing plate is
fractured, this value appears to be lower than the bearing strength predicted by existing design
formulae. The ultimate load was found to decrease as the starting position of the spiralreinforcement moved away from the upper surface. The ultimate load was found to decrease
when the outer diameter of the spiral reinforcements was decreased. This result showed atendency similar to that of the bearing strength calculation result of the PTI design equation.
With regard to damage to ribs and concrete voids due to defective anchorage device products, theultimate load was found to be 15% to 23% lower than in the design strength. A large difference
was observed depending on the product defects or presence of voids, and the difference in theultimate load depending on the size of defects or voids was small.
The Ultimate load of the model, which variables the voids of concrete, failure of the
bearing plate, and defects of the anchorage device, was derived below the ultimate strength
calculated through the bearing strength formula. Therefore, these variables are judged to be themain cause of failure in the anchorage zone during construction. To prevent the failure of theanchorage zone, it is important to design these components appropriately. In addition, during
construction, it is necessary to check whether the production of the anchorage device is abnormal,
and attention should be paid to the quality of the concrete construction materials to prevent voids
from forming.
5. Acknowledgements
This work was supported by the National Research Foundation of Korea(NRF) grant funded by
the Korea government(MSIT) (No. 2022R1F1A1074002).
6. References
EOTA. 2016. Post-Tensioning Kits for Prestressing of Structures (EAD 160004-00-0301),
Brussels, Belgium; European Organization for Technical Approvals (EOTA)
Kim, J. K., and Yang, J. M. 2021. Analysis of Bearing Strength of Post-Tensioning Anchorage
Zone with Respect to Relative Bearing Area and Lateral Confinement Design. Applied Sciences11(18), 8386.
Kim, J., Kwak, H. G., Kim, B. S., Kwon, Y., and Bouhjiti, E. M. 2019. Finite Element Analyses
and Design of Post-Tensioned Anchorage Zone in Ultra-High-Performance Concrete Beams.
Advances in Structural Engineering 22(2), 323-336.
Kwon, Y., Kim, J. K., and Yang, J. M. 2018. Development of Efficient Anchorage Device andEstimation of Its Bearing Strength of Posttensioning Anchorage Zone. Journal of StructuralEngineering 144(3), 1-10.

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Development of Shrinkage and Creep Prediction Model for SCC(B4-TW-SCC) and Implementation in Bridge Design Software
Wen-Cheng Liao1* and Jenn-Chuan Chern212: National Taiwan University, Taipei, Taiwan; email: wcliao@ntu.edu.tw: National Taiwan University, Taipei, Taiwan; email: jcchern@ntu.edu.tw*: corresponding author

Keywords: SCC; shrinkage; creep; prediction model
Abstract: At present, the design methodology and formula, particularly in shrinkage and creep,of Self-Compacting Concrete (SCC) projects in Taiwan are identical to those of Vibrated
Concrete (VC), but there are differences in the composition and proportion of the materials invarious regions, and obvious differences between the VC and SCC in engineering properties. In
this study, the test data of shrinkage and creep in Taiwan and abroad were collected following
the framework and parameters of the Northwestern University (NU) database to establish cloudbased analysis SCC database. Based on the Model B4-TW shrinkage and creep prediction model
modified by Model B4, this study revised the characteristics of SCC and the localization todevelop B4-TW-SCC. In B4-TW-SCC, the modification is mainly about the effect of highvolume of cementing material weight and high sand ratio, as well as the effect of drying
shrinkage changes on the drying creep. It is worth noting that efficiency in data mining andanalysis of huge sets of data are significantly improved by analysis database in development of
B4-TW-SCC. The creep curve and data point for specific SCC can be further generated on thiscloud-based analysis database and directly implemented in common bridge design software forpractical application.
1. Introduction
Creep and shrinkage of concrete, which are found under the influence of environmental effects
and conditions of long-term stress, affect the strength, deformation, and durability of structuralmembers, and in turn affect the life-cycle and serviceability of structures. In addition, long-spanstructures like bridges, high-rise buildings, and domes also require displacement control during
the construction phase in order to achieve the design criteria. Creep and shrinkage are influenced
by the materials and the mix design of the concrete, as well as surrounding stressors andenvironmental temperature and humidity. Experiments and prediction models should be used asa basis during the design phase of construction. Moreover, various types of high-performance
concrete, such as high-strength concrete, light-weight concrete, self-compacting concrete, fiber
reinforced concrete, fly ash or slag concrete, have been developed and taken into practical use
due to advances in material technology and the recycling of resources. When using new types ofconcrete, such as those mentioned above, experiments and prediction models are also required toclarify the properties of creep and shrinkage. However, experiments for determining those340

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properties of concrete are time-consuming and costly. In order to shorten construction time, nonmajor structures which are not deformation-sensitive usually rely on prediction models or the
combination of prediction models and short-term experiments.
Typical procedures for the development of prediction models are: (1) Collection ofexperimental creep and shrinkage data, (2) Analysis of existing prediction models, (3)Establishment of a database and analysis program, and (4) Establishment of a prediction modelof the target (or local) material. As result, major research institutions, including the ACI (1982),
CEB (1991), RILEM (2013), AASHTO (2014) and JSCE, academic institutions or scholars
including the University of Ottawa, Zdeněk Bažant of Northwestern University (1995), and JennChuan Chern of National Taiwan University (2018), had established databases or developed
prediction models respectively. However, the source, amount, and bases for comparison of datafrom each database are different, which often leads to discrepancies when comparing theperformance of prediction models. As a result, Bažant and others established large-scale
databases open for public use. Due to the continuous increase of research on concrete creep andshrinkage, the scale of databases has also continued to grow. Limited to traditional data fileformats and analytical methods, previous studies encountered many difficulties.
This study presents a new processing program for the use of concrete deformation databasesin Taiwan and extends to high performance concrete. This database is the first analysis systemand fast-access cloud database for shrinkage and creep of concrete in the world, named“shrinkage and creep database in Taiwan”, SCDT (http://scdt.ncree.org/). SCDT uses MS-SQL
as its database management system; it also uses Python programming language in the writing of
programs for analysis and post-processing. It has been found that these new processing programs
are far superior in terms of efficacy than previous types. In fact, when analyzing large amountsof data, they can be over several times as effective. In addition, per the needs of research,
convenient SQL syntax may be used to quickly perform selection comparison on the multiple
parameters which affect concrete deformation and their range, and thus easily perform crossanalysis, a process which is extremely time consuming in older methods.
2. Database Analysis Method of SCDTSDCT collects 1,412 sets of creep data and 1,995 sets of shrinkage data for vibrated concrete and475 sets of creep data and 1,587 sets of shrinkage data for self-compacting concrete from aroundthe world. Each set of data consists of data points taken at different times in a single creep or
shrinkage test. The purpose is to analyze and compare data among the TW, NU, and JSCE fromthe past 40-plus years. Besides the considerable scope of data, this study also faces challenges,
such as the variety of admixtures, range of concrete strength, type of cement, and data mining ofcomplicated conditions. As a result, the two traditional methods mentioned above are unable toprovide high levels of efficiency for this study. Instead, this study uses a new procedure for
database management and analysis—using Python and MS-SQL to integrate data access, mining,and high-efficiency analysis on a single platform.
SCDT used MS-SQL for the establishment of databases and Python for the programming ofprediction models. Taking advantage of the portability of SQL, this study successfully integrates
into a single interface the automatic output of data filtering, data mining, formulaic analysis,
graphical results, and statistical indicators. This new database management procedure, whichuses MS-SQL and Python. It may be seen that the inconveniences resulting from outputting andinputting data over different interfaces (inconvenience experiences in traditional databasemanagement procedures) have been greatly reduced, accompanied by an increase in the341

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efficiency of research in concrete deformation. By the use of an intuitive and easy mining
language, SCDT can accomplish analyses which would have been difficult or time-consuming
using the traditional methods of analysis, such as searching for data sets whose mix design
properties are similar to those in Taiwan’s national databases.
3. B4TW-SCC for Creep Prediction of SCC
B4TW-SCC was developed based on B4 model and optimized by data-mining with SCDT. The
first is the static modulus of elasticity which has different formulae among the available
prediction models for conventional concrete. To have highly accurate formulae of elastic
modulus for SCC, the relationship between this modulus and compressive strength should bedetermined initially before other optimization steps for creep compliance are carried out. Second,
the instantaneous compliance q1 was verified for different loading ages because this parameteranchors the entire multi-decade prediction model (Wendner et al., 2015a). In the next step, thebasic creep compliance parameters q2, q3, and q4 were optimized. The final step is calibration
for drying creep compliance parameter q5. In each step of optimization, the parameters
depending on the concrete composition were considered in the process, the water-to-cementitious
material ratio was found to be the most important for the magnitude of all the components of thecompliance parameters, and second most important is the aggregate to cementitious material
ratio (Wendner et al., 2015b). The individual influencing parameters were identified using
various statistical coefficients R2 and NRMSE, and screening assessments between predictedvalues and experimental data were used to have the highest efficiency.
The results B4TW-SCC model was optimized for the total creep function of SCC using
cement type R as illustrated in Fig. 1. All the dots were scattered uniformly around the standardline (x=y), and most of them were distributed in two red dash lines with an allowable error
range of ±40%. In addition, the red and blue dots in the plots represent the test data collectedfrom Taiwan and other countries around the world, respectively. For the purpose of estimating
the accuracy of the B4TW-SCC model, the evaluation was carried out based on the comparisonof this model and other prediction models for SCC as well as the models for conventional
concrete, including B4, fib MC2010 and ACI 209R-92. As the results are shown in Fig. 2, theB4TW-SCC model had the lowest values of statistical coefficient NRMSE, which indicated theminimum error in both cases of creep prediction for the NTU-SCC database.

(a) B4TW-SCC model(b) B4 model

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(d) ACI 209R-92 model(e) CEB90 (Poppe & Schutter)model(f) CEB90 (Kavanaugh) model(g) JSCE (Aslani) modelFig. 1. The predicted comparison in total creep of SCC
4. Example: Long-Term Deflection Analysis of Prestressed Bridges
In practice, when designing and evaluating the long-term deflection and service life of bridges in
Taiwan, international shrinkage and creep predictions models are often used. The morecommonly used prediction models include ACI, AASHTO and CEB-FIP. Whether the above
models are used to predict the shrinkage or creep of concrete in Taiwan, they are obviously
underestimated. If they are directly applied, the long-term deflection and prestress loss of bridges
are notably underestimated. This study selects an existing pre-stressed concrete bridge built by
the cantilever construction method, uses MIDAS Civil to build a model, and applies a variety ofshrinkage and creep prediction models, including ACI 209R-92, AASHTO LRFD 2014, CEB
MC90, CEB MC10, GL2000, Model B3 and Model B4TW (2020), etc., to discuss thedifferences in the analysis of different prediction models for bridge long-term deflection andprestress loss.
The case selected in this study is the Anshuo to Caopu section of the Taiwan Provincial
HighwayNo 9. This case has 3 spans in total, with a total length of 310m (80+150+80). For 69
segments of construction, the mix design and environmental parameter are listed in Table 1. The
loads considered in the analysis of this study include the self-weight, the weight of the partition
beam, the wet weight of the concrete, the load of the working vehicle and the formwork, the loadof the second phase and the pre-stressed steel tendons, etc.

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Table 1. Mix proportions and environmental parameters used in simulation
cementggbs
fly ashw/cm
agg/cm(kg/m3)(kg/m3)(kg/m3)0.40
3.55
356.2574.25
47.5
coarsefine aggregate
R.H.
V/S
aggregate
cement type
3
(kg/m
)
(%)
(cm)
(kg/m3)
977.3
707.7
80
27.45~43.71
I

f’c28
(MPa)
42.8
curing ages
(day)
3

There are obvious differences in the prediction results of different shrinkage and creep prediction
models in this case. Taking the segment with a V/S of 27.445cm and a loading age of 7 days as
an example, the total shrinkage prediction and total creep results are shown in Figure 2 (a) and(b), respectively. It can be found that the prediction results of ACI 209R-92 and AASHTOLRFD 2014, regardless of shrinkage and creep, have the phenomenon of rapid development in
the early stage and flattening at the earliest. This is an error caused by unreasonableconsideration of the influences of V/S. For other models, even though the development trend isrelatively close to that of Model B4TW (2020), the predicted values are very different. It can beseen that if it is applied to the analysis of prestressed bridges, the long-term deflection andprestress loss will be notably underestimated.

(a)

(b)

Fig. 2. The predicted comparison for different models for (a) total shrinkage and (b) total creep
The long-term deflection of prestressed concrete bridges can be composed of static load,prestressed tendon, shrinkage and creep. The total deflection and components of each deflection
at the middle of the bridge are shown in Figure 3. The deflection caused by static load has
nothing to do with shrinkage and creep, so the results obtained by all prediction models are thesame. The deflection of tendon arching will be affected by shrinkage and creep. The greater theshrinkage and creep, the greater the loss of the tendon. The deformation development of obvious,
shrinkage and creep is similar to the material shapes in Figure 2. From this case analysis, it canbe found that the greatest contribution to the long-term deflection of the bridge is creep, which
accounts for about 52% of the total deformation, and it develops rapidly in the early stage, andtends to be flat in about 2500 days; although the deformation caused by shrinkage only accounts
for the total deformation 14% of that, but it still continues to occur when the bridge age reaches
100 years, which is one of the main reason for the continuous deformation of the bridge.

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(a) Total deflection(b) deflection caused by dead load(c) deflection caused by tendon(b) deflection caused by shrinkage(b) deflection caused by creep

Fig. 3. Simulated deflection obtained by different prediction models
Since V/S is an independent variable for calculating shrinkage and creep, bridge segment withdifferent V/S will have different shrinkage and creep prediction values. In the above analysis,
each type of creep change needs to input 41 loading age curves, and each curve needs to input
108 points, and there are 19 V/S in this case, so a total of 41*108*19=84132 points need to be
input. Since the shrinkage has nothing to do with the loading age, it only needs to input
108*19=2052 points. After analysis in this study, it is found that if the whole bridge is analyzed
only by the average V/S, that is, all segments are calculated only by using the average V/S, thecomposition of each deflection analysis obtained in 100 years will be very close to the exact
analysis. Thus, this study suggests that the average V/S can be used in practice.

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5. Conclusions
This study presents the first analysis system and fast-access cloud database for shrinkage andcreep of concrete in the world, SCDT. Comparisons of the predicted results of selected models
and test results in the chosen database can be generated in seconds. One example of thedevelopment of B4TW-SCC based on model B4 by using SCDT to reflect concrete
characteristics in Taiwan and its implementation in bridge design is also presented in this study.
The following conclusions can be drawn:•••This study builds an online real-time analysis system S.C.D.T through cloud-based
databases and embedding Python into web pages created using ASP.net. Engineers andscholars around the world can use the three major functions provided by S.C.D.T without
installing any software or modifying codes, including shrinkage and creep prediction,
database analysis, and real-time analysis of test data upload.
This research uses a variety of shrinkage and creep prediction models in Taiwan and abroad,including ACI 209R-92, AASHTO LRFD 2014, CEB MC90, CEB MC90, GL2000, ModelB3 and Model B4TW (2020), etc., to evaluate the long-term performance of prestressed
concrete bridges. From the analysis results, it can be known that if the foreign prediction
model is directly applied without considering the characteristics of Taiwan concrete, thelong-term deflection and prestress loss will be seriously underestimated.
From this case, it can be found that creep deformation contributes the most to the long-term
deformation of prestressed concrete bridges, and the deformation caused by it accounts forabout 52% of the total deformation. Although the deformation is only 14% of the total
deformation, the bridge still continues to develop when it is 100 years old, which is themain cause of the continuous deformation of the bridge.

6. References
AASHTO (2014), AASHTO LRFD bridge design specifications 7th edition, American
Association of State Highway and Transportation Officials (AASHTO), Washington DC, USA.
ACI Committee 209 (1982), Prediction of creep, shrinkage and temperature effects in concrete
structures, Designing for Creep and Shrinkage in Concrete Structures, A Tribute to Adrian Pauw,
American Concrete Institute, Farmington Hills, MI, USA.
Bažant, Z., and Baweja, S. (1995), “Creep and shrinkage prediction model for analysis anddesign of concrete structures: Model B3,” Materials and Structures, 28, 357-365, 415-430, 488495. https://doi.org/10.1007/bf02486204
Bažant, Z., Hubler, M., and Wendner, R. (2015), “Model B4 for creep, drying shrinkage andautogenous shrinkage of normal and high-strength concretes with multi-decade applicability,”
RILEM Recommendation TC-242-MDC.
CEB Bulletin (1991), Evaluation of the time dependent properties of concrete, Bulletind’Information No. 199, Comite European du Beton/Federation Internationale de la Precontrainte,
Lausanne, Switzerland.
Chern, J-C., Liao, W-C., Liu, T-K., Chin, W-Y. (2018), “Development and analysis of creep andshrinkage database of concrete in Taiwan,” Structural Engineering, 33(1), 103-116. (in Chinese)346Bridge Engineering Institute Conference 2023 (BEI-2023)
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FIB, The International Federation for Structural Concrete (2013), Fib model code for concrete
structures 2010, Berlin, Germany.
Mehta, P. K. and Monteiro, P. J., (2006). “Concrete – Microstructure, Properties, and Materials”(3rd ed.). McGraw Hill.
Wendner, R., Hubler, M. H. and Bažant, Z. P., (2015a). “Optimization method, choice of form
and uncertainty quantiﬁcation of Model B4 using laboratory and multi-decade bridge databases.”
Materials and Structures, 48, 771-796.
Wendner, R., Hubler, M. H. and Bažant, Z. P., (2015b). “Statistical justiﬁcation of model B4 formulti-decade concrete creep using laboratory and bridge databases and comparisons to othermodels.” Materials and Structures, 48 (4), 815-833.

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A Data-Driven Method for Predicting the Temperature Response ofComposite Bridges Based on Deep Learning Technology
Yanjia Wang1*, Dong Yang2, and Francis T.K. Au3
1: The University of Hong Kong, Hong Kong, China; email: ansleyw@connect.hku.hk: Guangzhou University, Guangzhou, China; email: yangd@gzhu.edu.cn
3: The University of Hong Kong, Hong Kong, China; email: francis.au@hku.hk
2*: corresponding author

Keywords: composite bridge girder; deep learning; temperature behavior
Abstract: As climate change is imminent, it is necessary to investigate the possible effects on
bridges in future. Towards the end of this century, bridges may experience significant increase inthermal loading. The Structural Health Monitoring (SHM) systems installed on some bridges canprovide useful data to form the basis not only for verification of the design loading, but also for
estimation of the future scenarios. Recent developments in the field of Artificial Intelligence (AI)have opened up new opportunities. Compared with the traditional Finite Element (FE) Method toanalyze the structural behaviour under ambient conditions, Deep Learning (DL) as a data-driven
method offers an efficient alternative for simulating the temperature responses of structures
based on the massive real-time SHM data available. Long Short-Term Memory Networks(LSTM) and Gate Recurrent Unit (GRU) are two typical Recurrent Neural Network (RNN)
methods, which perform well in predicting the structural temperature and its effects. In this study,
the nonlinear relationship between the ambient climate condition and temperature response of abridge is predicted by the ambient condition input, with the bridge temperature sequential DL
models built separately based on LSTM and GRU with monitoring data from the temperature
sensors of a composite girder bridge. Comparison between the forecast structural temperature by
the LSTM and GRU model with the simulated results by the traditional FE Model confirms theexcellent performance of the proposed data-driven method.
1. Introduction
The impact of environmental loading on bridge structures tends to be increasingly significant due
to climate change. With increasing ambient temperature, the evaluation of bridges, especiallylong-span bridges, becomes more important. The challenge is how to predict the temperature
response of the bridge under the given climate conditions accurately. The method widely used byresearchers is analysis with Finite Element (FE) models. However, most of the FE analyses rely
on commercial software and are time-consuming. With the development of Machine Learningtechnologies, Deep Learning (DL) technology came to the stage and different kinds of DL
networks are used to deal with different scenarios. Recurrent Neural Network (RNN) is a kind of
Artificial Neural Network (ANN) using DL technology, which works well in solving timedependent problems. A proper choice of DL networks would likely apply to bridge temperature
response prediction. In this study, the prediction of the temperature response of a long-span

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bridge is conducted using two RNNs with different structures. The efficiency and performance ofthe networks will be compared.
2. Bridge Monitored
The bridge studied is Ting Kau Bridge (TKB), which is a 3-tower cable-stayed bridge in HongKong. Fig. 1(a) shows the composite bridge deck comprising concrete slabs supported on steel
plate girders. The data obtained by the Structural Health Monitoring (SHM) system of TKB
covered both the structural responses and ambient air temperature data. Temperature sensors are
installed on the main girders and cross girders on both sides of the symmetric deck as shown in
Fig. 1(b). From heat transfer theory, the three main factors that influence the bridge temperature
are the ambient air temperature, global solar radiation (GSR) I, and wind speed . In particular,
the ambient temperature includes the air temperature above the deckand that below the deck
. In this study, these four factors will be used as input to the proposed DL model and thebridge structural temperatures are the output. The hourly monitoring data of the year 2017 areused to train and validate the model.

(a)

(b)

Fig. 1. (a) Layout of the TKB composite deck; (b) Temperature sensors of the cross girder.
3. Methodologies
3.1. LSTM and GRU neural network
RNN employing DL technology is commonly adopted to deal with the sequential data of timeseries, such as translation, natural language processing, speech recognition, etc. (Lipton,
Berkowitz et al. 2015). As opposed to the basic Feedforward Neural Networks (FNN) andConvolutional Neural Networks (CNN), RNN takes information from the previous input together
with the current input to calculate the current output.
Long Short-Term Memory Networks (LSTM) (Hochreiter and Schmidhuber 1997) and
Gate Recurrent Units (GRU) (Chung, Gulcehre et al. 2014) are two main advanced RNNs, which
can effectively solve the gradient explosion or disappearance problem of basic RNN. Fig. 2
shows that the cell structures of the LSTM and GRU networks are controlled by Gating
Mechanism, and the functions with governing equations for each gate of LSTM and GRU
networks are given in Table 1 and Table 2, where b’s are the bias vectors, W’s are the weightmatrices, and σ is the activation function. In comparison, the structure of GRU is simpler, whichuses only an update gate to balance the input information and historical information without the
memory unit adopted in LSTM. Sufficient research has proved that GRU requires less training349

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time and less computational resources, but offers similar performance as LSTM (Fu, Zhang et al.
2016, Shewalkar, Nyavanandi et al. 2019).
Table 1. Gating mechanism of LSTM network
Gate name
Function
Determine the information to be“forgotten” from the internal state ofForget Gate
the previous moment

Governing equation(1)

Input Gate

Determine the information to be
accepted from the current candidate's
internal state

(2)
(3)
(4)

Output Gate

Determine the output from the current
internal state
to the external state

(5)
(6)

Table 2. Gating mechanism of GRU network
Gate name
Function
Determine the information to beretained from the previous state
Update Gateand the information to be accepted

Governing equation(7)(8)

from the current candidate state
Reset Gate

(9)
(10)

Determine whether the external stateshould rely on the previous state

(a)

(b)

Fig. 2. Cell structure of networks: (a) LSTM; and (b) GRU
3.2. Data preparation and Deep Learning DL model
A one-year monitoring dataset of TKB is used in this study. The data from Jan 1st to October
31st are used in the training process as training sets and validation sets. The remaining twomonth data from Nov 1st to Dec 30th are used as the test data to evaluate the prediction
performance by comparing the forecast results with the monitoring data. To enhance the qualityof training, the raw data is normalized before training. The performance of DL models is also
evaluated by comparing with the simulated results by the Finite Element Analysis (FEA) using
ABAQUS. For the DL models with LSTM and GRU networks, there are three hidden layers,

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with 256 cells for each layer. There is a fully connected dense layer linking the last LSTM or
GRU layer to the output layer. For the other hyperparameters, the Adam optimization algorithm
is used and the activation function for LSTM and GRU layers are Rectified Linear Unit (relu).
The length of the time series determining the data structure of the sequence is essential, which is
set to 24 hours, implying that the prediction of current structural temperatures is based on thehistorical measurements of the previous 24 hours. To evaluate the performance of the proposed
DL models, the Mean Absolute Error (MAE) to reflect the training loss and the accuracy
indicator
will be calculated in terms of the output data , the predicted value , the mean ofobserved output , the sum of squares of residuals
and the total sum of squares
as:
(11)
(12)
4. Results and conclusions
4.1. DL model performance
During the training process, the performance variations of the model as shown in Fig. 3(a)indicate that the training loss and accuracy of the LSTM and GRU models are close while the
GRU networks perform better in the prediction of structural temperature based on the error
calculated by the test datasets in Table 3. However, the training time for GRU networks is only
70% of the training time for LSTM networks according to the average training time for 10 times
as shown in Table 3, which proves that the GRU structure has higher training efficiency. Fig. 3(b)provides the statistical results of the temperature differences between the values predicted by DL
models and the monitored data. By comparing the predicted bridge temperature variations by the
LSTM and GRU models with the FE simulations in Fig. 3(c), the discrepancies of the predicted
results of both DL models are close to the discrepancy between the FE simulation and monitored
data. While it takes around 3 minutes to complete the FE analysis, the DL model only takes 1s tofinish the prediction, which can effectively achieve real-time prediction.

(a)

(b)

(c)

Fig. 3. (a) Loss and accuracy during training; (b) discrepancy between monitored and predicted
temperatures; (c) comparison of monitored data, DL model predictions and FE simulations

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Table 3. Training time and test error of LSTM and GRU networks
LSTM
Training time (s)2215
MAE of the test sets (deg)
0.551

GRU
3346
0.370

5. References
Chung, J., C. Gulcehre, K. Cho and Y. Bengio (2014). "Empirical evaluation of gated recurrent
neural networks on sequence modeling." arXiv preprint arXiv:1412.3555.
Fu, R., Z. Zhang and L. Li (2016). Using LSTM and GRU neural network methods for traffic
flow prediction. 2016 31st Youth Academic Annual Conference of Chinese Association ofAutomation (YAC).
Hochreiter, S. and J. Schmidhuber (1997). "Long Short-Term Memory." Neural Computation
9(8): 1735-1780.
Lipton, Z. C., J. Berkowitz and C. Elkan (2015). "A critical review of recurrent neural networks
for sequence learning." arXiv preprint arXiv:1506.00019.
Shewalkar, A., D. Nyavanandi and S. A. Ludwig (2019). "Performance Evaluation of Deep
Neural Networks Applied to Speech Recognition: RNN, LSTM and GRU." Journal of Artificial
Intelligence and Soft Computing Research 9(4): 235-245.

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Topology Optimization of Concrete Bridges
Prestressed with Shape Memory Alloys
Minsoo Sung1 and Bassem Andrawes2*
1
2: University of Illinois at Urbana-Champaign, Urbana, US; email: minsoos2@illinois.edu: University of Illinois at Urbana-Champaign, Urbana, US; email: andrawes@illinois.edu*: corresponding author

Keywords: topology optimization, prestressed concrete, shape memory alloys, bridges
Abstract: Topology optimization has been widely used as a design optimization tool that
distributes material at specific locations within the design domain to maximize or minimize
design objectives such as stiffness and/or maximum displacement. One of the main challenges ofmanufacturing topology-optimized concrete structures for real-world applications is theircomplex geometry. Complex concrete geometries are not only difficult to build but also hard tobe effectively reinforced with conventional steel reinforcement because the distribution of thetensile zones is not as simple as that of typical reinforced concrete designs. The key advantage ofusing shape memory alloys (SMAs) as prestressing reinforcement for topology-optimized
concrete bridges is that they can prestress target regions where only tensile stresses are expected.
SMAs have a unique thermo-mechanical characteristic, which allows them to recover their
original configuration from an excessively deformed state once heated. Using this characteristic,
SMA elements can apply prestressing forces wherever and whenever once they are triggered byheat, without any hydraulic jacking system involved. In this study, topology optimization of a
two-span bridge girder prestressed with SMA reinforcement is presented to prove the feasibilityof using SMAs to produce prestressed lightweight bridge members. During the topology
optimization, stress constraint is used to avoid high-stress concentration either from theprestressing or the external live loads. The results of the study prove the efficacy of SMA in
prestressing local regions in the topology-optimized bridge girder.
1. Introduction
Topology optimization (TO) is a powerful technique for optimizing the design of a structure byminimizing an objective function, such as compliance or maximum deflection, subject toconstraints like total material volume. TO results in a truss-like structure that is lightweight,
economical, and environmentally friendly. However, the implementation of TO in the real-world
construction industry is impeded by the complex geometry generated by the optimization process.
In particular, complex geometry in TO concrete members poses a manufacturing challenge andmakes it difficult to control tensile cracks. To prevent tensile cracks in concrete bridges, theconstruction industry uses pre- or post-tensioning techniques using high-strength steel (HSS)
reinforcements. However, with the conventional prestressing methods, it is inevitable to apply
prestressing to the entire region of the bridge, even at the regions where prestressing is notnecessary, such as compressive zones and end regions. The localized prestressing system using

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shape memory alloys (SMAs), which could apply prestressing only at the target region, wasproposed by (Sung & Andrawes, 2021) to overcome the limitations of the conventionalprestressing systems. The localized prestressing system is a suitable application for the TO
concrete member as it has a complex geometry with numerous localized tensile zones, whichconventional prestressing system could not effectively control the tensile cracks of the bridge. In
this study, the topology of the concrete and the SMA prestressing reinforcement are optimized
simultaneously in a commonly used bridge girder to come up with a lightweight truss-like design.
2. Localized Prestressing System Using Shape Memory Alloys
Shape Memory Alloys (SMAs) are a family of materials exhibiting a unique thermo-mechanicalcharacteristic called the shape memory effect (SME). The SMA wires that are extremely
deformed can recover their original configuration by activating the phase transformation of thematerial. The phase transformation occurs when the material is subjected to sufficient
stimulation energy, causing its microstructure to reorient. On a macroscopic level, if the material
is heated to its phase transformation activation temperature, the excessively deformed martensitic
alloy can recover its original form. However, if the material is restrained during heating,recovery stress is generated within the material, which is a source of prestressing force for SMA
prestressing applications. Sung & Andrawes, 2021 proposed a localized prestressing systemusing SMAs that could apply the prestressing effect only at target regions where tensile cracks
are expected. Pre-strained SMA reinforcements can be placed at the desired location before
casting concrete, and the prestressing force can be applied by activating the SMA reinforcements
once the concrete is fully cured. The new system could delay the application of prestressing force,
which could significantly reduce the prestress loss from the creep and shrinkage of the early-age
concrete. The new SMA-based prestressing system has the potential to reinforce complex local
tensile zones within topologically optimized concrete members, eliminating the need formechanical prestressing. To prove its feasibility, a study was conducted on a continuous-span
concrete bridge girder, considering real-world loading conditions as per current design guidelines.
3. Topology Optimization of Prestressed Concrete Bridge Girder
In this study, a two-span continuous girder was selected as an application example which was
designed to match the American Association of State Highway and Transportation Officials(AASHTO) Type-II girders with span lengths of 18.28m (Fig.1). The concrete was assumed tohave a compressive strength of 55.17MPa. The design load was calculated based on AASHTOLRFD specifications (AASHTO, 2020) under the HL-93 live load. The maximum front and rearaxle spacing of 4.27m was used for the HS-20 design truck to maximize the force effect (Fig.1).
In the study, the HS-20 truck was located at 9 different locations with 1.22m of intervals as anindependent load case. The live load distribution factors for multiple loaded lanes were
calculated based on AASHTO LRFD specifications. The HS-20 design truckload was adjustedwith a dynamic allowance factor of 33% and a multiple presence factor. The girder was modeledusing a hybrid approach with concrete elements modeled as 2D continuum elements and SMA
bars modeled as truss elements in MATLAB. Concrete elements were meshed with a 51mm-by51mm grid with the section width discretized to match the configuration of AASTHTO Type-II
girders (Fig.1). The width of the section was considered as the thickness of the 2D-girder model.
The SMA reinforcement was placed in layers and meshed using a 102mm-by-102mm grid withdiagonal elements considered (Fig.1). Both concrete and SMA elements used linear elastic

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material models with SMA reinforcement made of NiTiNb alloy, which exhibits a recovery
stress of 550MPa (Sung and Andrawes, 2021).

Fig. 1. Cross section and elevation view of the modeled
3.1. Topology Optimization framework
The TO framework was developed in MATLAB based on the work of Amir and Shakour, 2018with modifications for the SMA-based prestressing system. The developed framework wascomprised of two parts: FEM and Optimization module, as shown in the flowchart below (Fig.2).
In the FEM module, the system was modeled based on the design variables, and the systemequation is solved to get the displacement field (Fig. 2). The design variables (DVs) of the TO
are densities of concrete elements ( ρc ) and SMA elements ( ρt ), which are bound from zero to
one by the box constraints. The objective of the optimization was to minimize the amount of
deformation that occurs in areas where multiple external loads are applied while taking into
account the total volume constraints for both concrete and SMA elements. The deformation ofeach external load case i ( φi ) was calculated using the equation shown in Fig. 2, which was
similar to finding the compliance of the system, where the external force vector for the load case
I is fexti , and the total displacement vector is utotal . To prevent negative values, the deformationof each load case was squared. To take the worst load case into account in objective function
evaluation, the Kreisselmeier-Steinhauser (KS) function was utilized to calculate the maximumof φi for multiple load cases, as described in Fig.2. The KS function was chosen due to its
differentiability, which was essential for the optimization process based on gradient descent. The
DVs were filtered with density, embedding, and Heaviside filters to ensure the SMA
reinforcements were embedded in the surrounding concrete and to avoid checkerboard patterns.
The solid isotropic material penalization (SIMP) method was used for the elasticity moduli of theconcrete and SMA elements.
The optimization module is comprised of three components: evaluating the objective function,
conducting sensitivity analysis, and using the MATLAB function “fmincon” as the optimizer.
The objective function is evaluated based on the solution obtained from the FEM analysis, andboth the objective and constraint functions underwent sensitivity analysis. The “fmincon”
optimizer updates the DVs using gradients calculated from previous iterations. The use of anaugmented objective function and the method of adjoint vectors allows for efficient sensitivity
analysis. The differentiation of the augmented objective function results in the calculation of theadjoint vector and sensitivity of the i-th objective function ( φi ) with respect to the DVs.

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Input Constraints( , )Optimization module

Initialize DVs
(
,
)

No

•

Evaluate Objective Function
For each , total deformation squared induced by•

Maximum of

Sensitivity Analysis
Objective sensitivity;
Constraint sensitivity;

Output DVs

•
•
•
•

fmincon (MATLAB)
Optimality check
Update DVs ( , )End;

Optimum?
Yes

FEM module•••

2D FEM modeling
Filter DVs ( , )Evaluate Stiffness Matrix; SIMP (Evaluate force vectors (•Solve System Equations
Multi vehicle load cases;))(KS function);Fig. 2. Flow chart of the topology optimization framework
3.2. Optimization Results
As an example, the concrete volumetric ratio ( Vc* ) of 75% and the SMA reinforcement ratio( ρ SMA = Vt * / Vc* ) of 0.35% case were selected to show the optimization result (Fig.3). The
optimum solution showed a concrete topology and SMA reinforcement profiles which can bedivided into three regions: negative moment, positive moment, and end regions. The negative
moment region has SMAs located at the top chord, while the positive moment dominant region
has SMAs located at the bottom chord. The end region was heavily prestressed with top
reinforcements, which could contribute to uplifting the positive moment region. The
optimization results showed that the diagonal reinforcements at the transitional zones near the
5m to 16m sections helped in reducing the deflection at the positive moment dominant region.
The optimized girder design satisfied the stress limits as per AASHTO LRFD service I limit
states.
(4) SMA wires(2) SMA wires
Negative M(8) SMA wires
Positive M

End

Fig. 3. A half-elevation view of the topology-optimized girder with a concrete volumetric ratioof Vc* =75%, and 0.35%
4. Conclusion
This study presented a topology optimization framework for optimizing the design of acontinuous prestressed concrete bridge girder using SMAs. The optimized design resulted in
25% reduction in concrete volume compared to conventional designs and satisfied AASHTOLRFD stress limits. The study highlights the potential of using SMAs in topology-optimized
concrete structures.

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5. ReferencesSung, M., & Andrawes, B. (2021). Innovative local prestressing system for concrete crossties
using shape memory alloys. Engineering Structures, 247, 113048.
https://doi.org/10.1016/j.engstruct.2021.113048
Amir, O., & Shakour, E. (2018). Simultaneous shape and topology optimization of prestressed
concrete beams. Structural and Multidisciplinary Optimization, 57(5), 1831-1843.
American Association of State Highway and Transportation Officials (AASHTO) (2020).
AASHTO LRFD Bridge Design Specifications, 9th.

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A Mesoscale Probabilistic Fatigue Crack Initiation Model for Steel Bridges
Based on the Modified Fine and Bhat Theory
Fei Jiang1*, You-Liang Ding2, Kang Yang3, and Xiao-Nan Zhang4
1: Southeast University, Nanjing, China; email: Feijiang1995@outlook.com
2: Southeast University, Nanjing, China; email: civilchina@hotmail.com
3: Southeast University, Nanjing, China; email: 230189498@seu.edu.cn
4: Southeast University, Nanjing, China; email: nanan6677@163.com*: corresponding author
Keywords: Steel bridges; Mesoscale modeling; Crack initiation; Non-deterministic prediction
Abstract: The safety and durability of steel bridges are strongly limited by the deterioration
process of fatigue. The fatigue initiation process accounts for a large proportion of fatigue life.
Therefore, performing accurate initiation life prediction can facilitate the fatigue maintenance ofsuch structures. In this study, Fine and Bhat theory was modified to build a mesoscale small
crack initiation model, which describes the crack nucleation in grains and the coalescence andpropagation of microcracks. A crystal plastic finite element model was established for obtaining
the plastic strain energy density parameter for the proposed model. To consider the
microstructure uncertainties, the grain size variation was included through inverse transform
sampling on the tested grain diameter distribution. Before prediction, the initiation model wascalibrated using prescribed historical fatigue data, and Monte Carlo simulation was thenconducted to predict the small crack initiation life. The proposed model was validated by
performing fatigue testing on full-scale segmental orthotropic steel deck specimens with typicalU-rib to diaphragm welded joints.
1. Introduction
The safety, functionality, and durability of steel bridge structures are strongly limited by fatigue
deterioration problems. As the increase of external loads and service time, there is a mounting
risk of bridge collapse within the designed service life. To provide a reference for the fatigue
maintenance activities, it is necessary to perform accurate life predictions for fatigue-vulnerable
details.
In fatigue life prediction tasks, the small crack initiation has received much attention since ittakes up a large portion of the total fatigue life. Phenomenological models, such as Basquin’s
model (Basquin, 1910) and Manson-Coffin’s model (Coffin, 1956), are still the main approaches
for fatigue analysis. However, these models are based on extensive fatigue experiments, which iscostly and time-consuming. Besides, macro fatigue limit, material microstructure effect, andnotch effect are often sacrificed, which may introduce significant error in some circumstances(Schijve, 2009; Yuan, Zhang, Kim, & Liu, 2017).
In recent years, the mesoscale modeling of the small crack initiation stage has gained
popularity. Castelluccio et al. (2014) presented a thorough review on advances in assessing

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microstructure-sensitive early-stage fatigue of polycrystals and proposed that further efforts stillneed be paid on probabilistic methodologies to quantify uncertainty in predicting fatigue
initiation life. This study proposed a mesoscale probabilistic fatigue crack initiation model,
which aims to accurately predict the fatigue initiation life for steel bridges considering thematerial microstructure attributes and uncertainties.
2. Fatigue Experiment
An application example based on orthotropic steel deck (OSD) structures was employed todemonstrate the proposed model. The fatigue initiation at the typical U-rib to diaphragm welds
was monitored by performing fatigue testing on a full-scale segmental OSD specimen. The
specimen and test setup are illustrated in Fig. 1. The specimen consisted of an 18-mm-thick deckplate, two 8-mm-thick longitudinal U-ribs, and two 14-mm-thick transverse diaphragms. The
global length, width, and height of the specimens were 2,900, 1,200, and 594 mm, respectively.
All the steel components were made of Chinese steel type Q345. The U-rib and diaphragm wereconnected through fillet welds.
The OSD specimen was fixed to a testing platform by applying a cyclic pressure load withthe maximum value of −20 kN, the minimum value of −260 kN, and the loading frequency of 4
Hz. A rigid load spreader was employed to transfer the concentrated actuator force into an
approximate uniform area load. To avoid direct rigid contact between the spreader and thespecimen, a rubber layer with dimensions of 400 × 400 × 54 mm was positioned between the
spreader and the deck plate to simulate a single wheel-tire contact area. During the fatigue testing,strain gauges were adopted to monitor the local fatigue stresses, and the fatigue initiation timing
was recorded with the assistance of a hand-held electronic magnifier.

Fig. 1. Fatigue experiment on a full-scale segmental OSD specimen
3. Mesoscale Modeling of Small Crack Initiation
3.1. Modified Fine and Bhat theory
From an energy point of view, Fine and Bhat (Fine & Bhat, 2007) hold an energy barrier must be
overcome for forming new surfaces during fatigue crack nucleation. It is presumed that theformation of new surfaces will result in a release of elastic energy from the applied load andaccumulated lattice defects. By differentiating the change of Gibbs free energy ∆G due to crack
nucleation with respect to the radius a of the half-penny-shaped virtual crack surface, therelationship between nucleation cycle N* and the critical crack size a* can be finally derived
after designating d∆G/da equal to zero:359

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N* = 4π Eγ s − 2σ 2 a* (1 −ν 2 )(1)

π Ef *ϖwhere E is Young’s modulus; ν is Poisson’s ratio; σ is the fatigue stress range; γs is the surface
energy of the material; f* = ftm; f is an energy efficiency factor, describing the portion of stored
external work due to slip irreversibility; tm is the maximum width of the persistent slip bands; ϖ= ∆τ∆γp/2 is the plastic strain energy density (PSED); ∆τ is the shear stress range; ∆γp is theplastic shear strain range. In this study, material constants E and ν for the steel Q345 were 210
GPa and 0.3, respectively. An estimated value of 2.4 J/m2 was adopted for γs of steel Q345 atroom temperature.
However, the Fine and Bhat theory only describes the crack formation in a single grain, andit does not account for the propagation and coalescence of microcracks. In this study, the Fine
and Bhat theory is was modified and extended based on several presumptions to provide
information about the small crack initiation life and its corresponding size. It is presumed that (1)the crack formation in every single grain follows the Fine and Bhat theory; (2) the stressconcentration at grain boundaries caused by crack formation in the previous grain is the trigger
for the start of slip systems in its neighboring grains; (3) the parameters γs and f* remain constantfor the crack formation in different grains of a given material; and (4) during cyclic loading,existing single grain cracks coalesce into a small crack, which is instantaneous and does notconsume the number of cycles. Fig. 2 presents the small crack initiation process described by themodified theory. The crack surface forms first in grain 1, then in grain 2, and so on. Growth
paths are set from grain 1 to the crack front of the small crack after coalescence. Therefore, thefatigue initiation life can be calculated as:n

N in = ∑ 4i =1π Eγ s − σ 2 di (1 −ν 2 )(2)

π Ef *ϖwhere di corresponds to the diameter of the ith grain in the growth path. Nin is linearly related tothe total path length Σdi from Eq. (8), so the initiated small crack surface is half-penny-shaped,
which is consistent with the current assumptions about the short crack shape for steel bridges.

Fig. 2 Small crack initiation process describedby the modified Fine and Bhat theory

360

Fig. 3 Illustration of the CPFE simulation
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3.2. Crystal plasticity finite element model
To obtain the PSED ϖ, crystal plasticity finite element (CPFE) simulation was conductedbased on the information provided by electron backscatter diffraction (EBSD) testing. A
representative volume element (RVE) model for microstructure at the fatigue initiation point wasestablished by Voronoi tessellation, as illustrated in Fig. 3. The RVE model had a length of 80µm and contained 512 grains according to the measured average grain diameter. A periodic
displacement was applied on the Y surface to realize the repetitive deformation at the crack
initiation position.
4. Probabilistic Prediction of Fatigue Initiation Life
4.1. Calibration of small crack initiation model
The small crack initiation process is non-deterministic due to the microstructural uncertainties ofngrain size variations. To consider this, the term ∑ di was included in Eq. (2). The parameter f*
i =1should be accordingly treated as a random variable F* and calibrated before life prediction.
Inverse transform sampling was performed on the grain diameters d along the surface crack
growth path according to the EBSD testing results. During the kth sampling, a sequence of grainndiameters {di }i =1 was generated. This randomly sampled diameter sequence was further filtered

(

)

k

given the constraint of the handshake small crack length, as described by the inequality (3).
d1 + 2∑ i = 2 di − a ≤ Ean(3)

where a is the handshake surface crack length and equal to 0.5 mm in this study; Ea is thenallowable sampling error, which was set to 1 µm. By delivering the sample {di }i =1 along with

(

)

k

historical initiation (σ, Nin) into Eq. (2), the kth realization f*k of F* was obtained. The sampling

process was repeated, and 12,451 samples were finally produced. The frequency density fitted by
the lognormal distribution was obtained, as shown in Fig. 4. During the experiment, the fatigue
initiation of two U-rib to diaphragm welds designated as D1 and D2 was recorded. The result (σ= 298.2 MPa, Nin = 110,000 cycles) of D2 was herein used as the historical data for thecalibration, while the result (σ = 340.8 MPa, Nin = 70,000 cycles) of D1 was employed tovalidate the life prediction results as described below.
65

Fitted PDF

Probability density (×10-5)

Frequency density (×104)54µ = −9.4702
σ = 0.09554

3

R2 = 0.99807
2
1
0

6

7

8

9

10

11

f* (×10-5 mm)

Fig. 4 Probability distribution of thecalibrated F*

4

3

2

1

0

12

Observed initiation life
Sample mean
Predicted initiation life

3

4

5

6

7

8

9

Fatigue initiation life (×104 cycle)

1011Fig. 5 Probability distribution of the predictedcrack initiation life of welded joint D1

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4.2. Probabilistic prediction of fatigue initiation life
After obtaining the calibrated F* distribution, Eq. (2) was well-equipped to predict the small
crack initiation life using Monte Carlo simulation (MCS). In this study, the fatigue initiation life
at welded joint D1 was predicted using the proposed method. Based on processes similar to thencalibration stage, a sequence of grain diameters {di }i =1 and f*k were generated during the kth

(

)

k

sampling. These generated parameters were submitted into Eq. (2) to output the kth sample of thepredicted initiation life Nink. Finally, the life distribution was obtained, as shown in Fig. 5. It canbe observed that the mean of the distribution is very to the experimentally observed life result.
This demonstrates the validity of the proposed method.
5. Concluding Remarks
This study proposed a novel method for mesoscale modeling of the small crack initiation life of
welded joints in steel bridges. The Fine and Bhat theory was modified to further account for the
microcrack propagation and coalescence during the fatigue initiation stage. Based on themodified theory, the initiated small crack surface was half-penny-shaped, which is consistent
with the current assumptions about the short crack shape for steel bridges. To consider the
influence of microstructure uncertainties, the model parameters were treated as random variables,
and the model calibration and life prediction methods were outlined. The validity of the proposed
method has been demonstrated by fatigue testing on a full-scale segmental OSD specimen.
6. References
Basquin, O. H. 1910. The exponential law of endurance tests. American Society for Testing andMaterials Proceedings, 625–630.
Castelluccio, G. M., Musinski, W. D., and McDowell, D. L. 2014. Recent developments inassessing microstructure-sensitive early stage fatigue of polycrystals. Current Opinion in Solid
State and Materials Science, 18(4), 180–187.
Coffin, L. 1956. Design aspects of high-temperature fatigue with particular reference to thermal
stresses. Transaction of the ASME, 78, 527–532.
Fine, M. E., and Bhat, S. P. 2007. A model of fatigue crack nucleation in single crystal iron andcopper. Materials Science and Engineering: A, 468–470, 64–69.
Schijve, J. 2009. Fatigue of structures and materials (2nd ed.). Dordrecht: Springer.
Yuan, H., Zhang, W., Kim, J., and Liu, Y. 2017. A nonlinear grain-based fatigue damage model
for civil infrastructure under variable amplitude loads. International Journal of Fatigue, 104,
389–396.

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Tension Estimation Method for Linked Suspenders Based on
Identified Mode-shapes
Chien-Chou Chen1*, Wen-Hwa Wu2, and Yen-Chih Peng3
1: National Yunlin University of Science and Technology, Douliu and Ynlin;email: ccchen@yuntech.edu.tw
2: National Yunlin University of Science and Technology, Douliu and Ynlin;email: wuwh@yuntech.edu.tw
3: National Yunlin University of Science and Technology, Douliu and Ynlin;email: a0931256616@gmail.com
Keywords: tension estimation; linked suspenders; mode shape; effective vibration length
Abstract: To release the effect of complicated boundary condition on determining accurate cableforce, an innovative approach based on mode shape was proposed recently by the authors. With
multiple measurements to identify modal frequencies and mode-shape ratios, it has been verified
numerically and then experimentally that excellent accuracy in tension estimation can beachieved with precise fitting of half-wavelength for the sinusoidal component of each chosenmode shape. This study aims to explore the feasibility of this approach in determining thetension of linked suspenders of arch bridges. It is confirmed with finite element method that themode shapes in each segment of the linked strands still mostly follow a sinusoidal function witha variation only observable near the boundary or linkage due to the influence of hyperbolic
components. The numerical analysis results show that accurate tension can be obtained with thisapproach. Then, the mock-up test with linked strands in laboratory is further conducted to verify
the applicability of the approach in practice. The experimental results demonstrate that the errorsof computed tension are pretty small. It is particularly noteworthy that the measurements
together with careful mode selection and appropriate choice in covering range of measurement
are all critical to hold such a superb accuracy.
1. Introduction
For efficiently tackling the difficulties associated with complicated boundary constraints, aninnovative concept combining mode shape functions and modal frequencies was recently
proposed by the authors to develop an accurate tension estimation method. The key of thismethod is to collect multiple synchronized vibration measurements on a tension member andthen identify the mode shape values at different sensor locations. The corresponding effective
vibration length for each considered mode can then be independently obtained by best fitting thesinusoidal component of mode shape. Previous works have indicated that the installation of
sensors cannot be concentrated near one end (Chen, 2018). Otherwise, it would be difficult toobtain reasonable effective vibration lengths through optimal fitting. Such a measurement
requirement, however, could greatly hinder popular applications of this method because a mobile
crane is indispensable for the sensor installation close to the high end in the case of stay cable orsuspender. Therefore, this research group has newly developed a novel approach for convenient363

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cable tension estimation simply based on local vibration measurement to fit partial mode shape.
It has also been verified that excellent accuracy in tension estimation can be achieved with
precise fitting of half-wavelength for the sinusoidal component of each mode shape when 1/3 ormore limited coverage range is adopted (Chen, 2022).
New concept of this method is based on the consideration of the mode shape function of atension member ignoring sag effect is the combination of a sinusoidal component and a
hyperbolic sinusoidal component (Chen, 2018). When zero-crossings of the sinusoidal
component are identified, the frequencies of the tension member can be explicitly expressed as
Eq. (1).
T+
f k2 =

π 2 EI( Lk / k ) 2
4m ( Lk / k ) 2(1)

in which f k represents the natural frequency of the mode with k half wavelengths in Hz, m
denotes the mass per unit length, E is the Young’s modulus, I stands for the cross-sectional area
moment of inertia, Lk/k is the half wavelength of the sinusoidal component and Lk is named as
the effective vibration length of considered range. Eq. (1) is similar with the frequency equationof simple supported beam except that the half wavelength of the sinusoidal component is used toreplace beam length. Accordingly, Eq. (1) can be applied to the tension member with more
complicated boundary constraints. Nonetheless, more complicated multiple measurements areneeded for identifying mode shape values to fit the sinusoidal component. To best fit theidentified mode shape values at n measurement locations for each considered mode, thecorresponding error function has been suggested as (Chen, 2022):()

 kπ x j + d k  ˆ 
e k ≡ ∑ ak sin  − φ jk 
Lk
j =1



n 

2

(2)

where ak and d k denote the amplitude coefficient and shifting parameter of the mode with khalf wavelengths, respectively, and φˆ jk represents the identified mode shape value at x j .
Because there is no need to consider boundary condition, it is believed that this method could beapplied to more complicated tension members in addition to single and uniform cable orsuspender. Linked suspenders are, therefore, selected to be the target for the further study in thispaper.
2. Numerical Exploration
The numerical analysis is performed to explore the applicability and accuracy of the method
based on mode shape applied on linked-cable system. Two linked strands chosen for conducting
the following mock-up test in laboratory are simulated in the finite element analysis with
SAP2000 software. The analysis model is shown in Fig.1 and then the modal analysis can be
executed to find the modal frequencies and mode shapes that serve as identified modalparameters. Five cases with different combinations of cable force and flexural rigidity, in which364

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hyperbolic sinusoidal components have different levels of contribution on mode shapes, arearranged for discovering the modal characteristics of linked-strand system. The results from themodal analysis reveal two anticipated features of modal parameters. First, different from single
strand with nearly arithmetic-sequence modal frequencies, linked-strand possesses irregularmodal frequencies. Second, the mode shapes in each sub-system, which is divided with thelinked point, still present similar sinusoidal function except in the neighborhood of linked point
and model boundary, as shown in Fig. 2, which is essential for the application of this method.

Fig. 1. Finite element analysis modelFig. 2. Mode shape of linked system

With the prepared FE models, five sensors, either concentrated at two ends or with uniform
spacing, are deployed within five considered converge ranges, 0.5ls, 0.6ls, 0.7ls, 0.8ls and 0.9ls. ls
represents the length of each sub-system. For the two-end deployment, five sensors are divided
into two groups, two sensors at one end and three sensors at the other ends. The numerical modeshape values corresponding to each sensor location are then optimally fitted to obtain theeffective vibration length for each mode based on Eq. (2). To more systematically examine theaccuracy of effective vibration length obtained from different sensor deployments, thecorresponding exact solutions are attained from Eq. (1) and taken as the reference values. Due tothe limitation of page number, only one mode shape from out-of-plane measurement and one
mode shape from in-plane measurement of the case 1, in which two strands have same tension
and flexural rigidity, are demonstrated in this paper. By comparing with the corresponding exact
values solved from Eq. (1), the error percentages of the optimal effective vibration length for
longer strand are computed and then plotted in Fig. 3 for in-plane measurement and Fig. 4 forout-of-plane measurement. Blue and green dots represent the error percentage of the calculated
effective vibration length for uniform deployment and deployment concentrated at two ends,
individually. Two red lines stand for the threshold of ± 0.5% relative error, which can also bedeemed ± 1.0% relative error of strand tension. From Figs. 3 and 4, it is clear to observe that theerror percentages from the deployment concentrated at two ends are more stable than those fromthe uniform deployment and increase gradually with coverage range until 0.9ls where it is about
0.5%. As for uniform deployment, there is one jumped error displayed at 0.9ls in Fig. 3 and one
jumped error at 0.8ls in Fig. 4, which are caused by the peculiar sensor deployment, all sensors
near the zero crossings of mode shape. For the other four cases, similar results are also observed.
3. Experimental Verification
Referring to the analysis results of the finite element model, mock-up test with two prestressed 7wire strands of 12.7 mm in diameter linked together and arranged as the drawing shown in Fig. 1,
is conducted in laboratory for further assessing the applicability of the method based on modeshape applied on linked suspenders in practice. MENS capacity accelerometers SDI2210-002
fabricated by Silicon Design Inc. were used to measure the ambient vibration signals of365

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experimental linked strands. 300 seconds of duration was taken for each measurement in theexperiment with a sampling rate 200Hz for longer strand and 500 Hz for shorter strand,
individually. Similar to the simulation cases in numerical analysis, three cases with different
strand tensions, which took the effect of hyperbolic component into account, were verified. Bothin-plane and out-of-plane vibration signals were measured for each case. Five sensors
concentrated at two ends of the measured strand was deployed within several chosen coverage
ranges. To identify the modal frequencies and mode shape ratios of the experimental strands,
discrete Fourier transform (DFT) was performed on the synchronized measurements. With theidentified mode shape values, optimal process was further implemented with Eq. (2) for
determining the effective vibration length of each mode. Then, the optimal value of strand
tension was computed with three selected modes. Since no precise values of strand tension andflexural rigidity are available to calculate the exact effective vibration length as in the numericalexamples, instead of effective vibration length, the strand tension determined from themeasurements of independent strands with full coverage concentrated at two ends were adopted
as the reference values in the experimental study.

Fig. 3. Deviation of effective vibration length(out-of-plane measurement)

Fig. 4. Deviation of effective vibration length(in-plane measurement)

Based on the numerical investigation in the previous section, the effective vibration lengths
obtained from two useless fitting curves of mode shape, coverage range less than one half
wavelength or the group with two sensors located at the peak or valley of the mode shape, are
not taken for the calculation of strand tension. In addition, the effective vibration length attained
from the fitting mode shape with larger error is also not considered. In comparison with thereference values, two important points are observed from the analysis results of three consideredcases. First, the deviation of the strand tensions calculated from in-plane measurement is muchworse than that from out-of-plane measurement, that could be caused by the damping effect of
the rubber pad of linkage device. Second, the largest deviation of the computed tensions of two
strands is approximately 2%. Although the results from the numerical study and mock-up test
validate that the method based on mode shape can estimate high accuracy tension of linked
strand, it is particularly noteworthy that the measurements together with careful mode selection
and appropriate choice in covering range of measurement are all critical to hold such a superb
accuracy for the practical application on the linked suspender of arch bridges.

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4. References
Chen, C.C., Wu, W.H., Chen, S.Y., Lai, G.L. 2018, A novel tension estimation approach forelastic cables by elimination of complex boundary condition effects employing mode shapefunctions, Engineering Structures, 166 , 152-166.
Chen, C.C., Wu, W.H., Liu, Y.T. and Lai, G.L. 2022, A Convenient Cable Tension Estimation
Method Simply Based on Local Vibration Measurements to Fit Partial Mode Shapes,
Engineering Structures, 272, Paper No. 1150081.

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Research on Crack Damage Model and Size Effect of Flexible Fiber
Reinforced Recycled Concrete
Yongcheng Ji1, Yanmin Jia1, Dayang Wang1*, Yanwei Jia1, and Zhiyang Pei1
1

: College of Civil Engineering and Transportation, Northeast Forestry University, Harbin
150040, China; email: yongchengji@126.com*: corresponding author

Keywords: Recycled concrete, Flexible fiber, Microscopic simulation, Compressive strength,
Size effect
Abstract: A mesoscale numerical model is presented to study the fracture damage mechanism offlexible fiber-reinforced brick aggregate recycled concrete (FFF-BAC). The model considers that
cement mortar matrix and flexible fiber are combined into a homogeneous material, namely
flexible fiber reinforced cement mortar matrix (FF-MC), and FF-MC is combined with coarse
aggregate and interfacial transition zone to form fiber reinforced recycled concrete. Combinedwith the experimental results of FF-MC in the research literature, the mechanical constitutive
model of FF-MC is determined.The numerical model of FFF-BAC was established, and theeffect of brick aggregate (BA) content and size on the mechanical properties of FFF-BAC isstudied based on the new model.
1. Introduction
Many scholars use the method of mesoscopic numerical simulation to study the reinforced
concrete with large stiffness fiber. Sun et al.(2019) studied the compression, splitting andbending properties of basalt fiber concrete through mesoscopic numerical simulation
experiments.However, the research on the numerical simulation of flexible fiber reinforced
concrete is lacking. This is because the flexible fiber itself is very thin, and millions or even tensof millions of fibers are needed in an ordinary recycled concrete test block, and it is difficult tosimulate the flexible fiber reinforced recycled concrete with the current computer level.
China specification GB50242019 has stipulated the conversion factor between nonstandard concrete cube and standard concrete cube. However, there are few researches on thesize effect of recycled concrete. Zhang et al. (2019) studied the effect of size effect oncompressive strength of recycled concrete with waste concrete aggregate with different
substitution rates. The higher the replacement rate of recycled aggregate, the lower the effect ofsize effect. Su et al. (2022) studied the size effect of the compressive strength of recycledconcrete, and found that the size effect showed an increasing trend with the increase of thereplacement rate of recycled aggregate, and the size effect degree was about 1.4 times that ofordinary concrete at the replacement rate of 100% recycled aggregate. In summary, whether thesize effect of recycled concrete is enhanced with the increase of recycled aggregate content. The
authors believe that the inconsistent conclusions are related to the quality of recycled aggregate.

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This article proposes a numerical model to study FFF-BAC, and based on this model, the size
effect of FFF-BAC is studied.
2. Experimental Program
2.1. Compression constitutive of fiber reinforced cement mortar matrix
In this paper, Zhou(2015) proposed a stress-strain relationship model for FF-MC undercompression. The stress-strain relationship of the ascending section can be expressed as:
E ε
 E0ε (1 − α )σ = 0=α(0 < ε < ε 0.4 )(ε 0.4 < ε < ε 0 )

aE0ε
−b
f cr'

(1)
(2)

Among them, a and b are 0.308 and 0.124, respectively. Considering the post-peak localization
characteristic of ECC, the stress-deformation curve is introduced to describe the post-peak
behavior of ECC. Bilinear curves are described as follows:
xl and σl are the deformation and stress at the inflection point of the softening curve, and x0 and
fcr’ are the deformation and stress at the peak load. That is, the stress and deformation at the peakload, the stress and deformation at the inflection point, and the stress and deformation at any
point on the curve. Statistical analysis of the test results shows that xl and σl are xl=1.5x0 and
σl=0.5 fcr’, respectively.

σ= m( x − x0 ) + f cr''σ = n( x − x0 ) + f cr

( x0 < x < xl )( xl < x < xmax )(3)

2.2. Compression constitutive of fiber reinforced cement mortar matrix
In this paper, a simplified bilinear hardening model proposed by Li (2009) is adopted. The
structure consists of two stages, namely, no damage stage and stable damage stage. See equation(4) for the tension constitutive formula.

(ε ≤ ε nt )E εσ =  nt
(ε nt < ε ≤ ε tu )
σ nt + Etu (ε − ε nt )(4)

1 σ nt;Etu is the elastic modulus of hardening stage. Ent is the elastic modulus of×
10 ε nt
the elastic phase. σnt is the cracking stress, εnt is the cracking strain, σtu is the ultimate stress, εtu isthe ultimate strain.

Where: E=
nt

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Bridge Engineering Institute Conference 2023 (BEI-2023)
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2.3. Fiber reinforced cement mortar matrix
Chen conducted compressive strength test and splitting tensile strength test of PVA reinforced
mortar. The detailed test process is shown in reference (Chen, 2019). The mechanical indexes of
FF-MC are shown in Table 1.
Table 1. FF-MC mechanical indicators
Compressive strength Modulus of elasticity
Group
（MPa）
（MPa）
FF-CM
42.36
24600

Poisson's ratio
0.206

Splitting tensile
strength（MPa）
6.39

The mechanical index of Table 2 is put into formula (1) - formula (3) to draw the compression
constitutive model of fiber mortar, as shown in Fig. 1(a). Bring the formula in Table 3 into
formula (4) to draw the tension constitutive model of FF-MC, as shown in Fig. 1(b).
6
5stress（MPa）

Stress（MPa）

40

30

20

10

0
0.000

4
3
2
1

0.003

0.006

0.009

0.012

0.015

0
0.000

0.002

strain

(a)

0.004

0.006

0.008

strain(b)

Fig. 1. (a)tension constitutive,(b) compression constitutive Fig. 2. Micro numerical model
3. Numerical Simulation
3.1. Numerical model of random aggregate
In this paper, ABAQUS was used to establish a random aggregate model of recycled concrete.
The typical model diagram is shown in Fig. 2. The volume fraction of aggregate is about 45%
The thickness of ITZ is set to 0.5mm. The matrix of fiber and cement mortar is regarded as ahomogeneous material. Three node planar stress triangle element (CPS3) is used in the gridelement.
3.2. Material parameter
The mechanical behavior of cement mortar and ITZ is similar to that of concrete, so the concrete
Plastic damage model (CDP model) in ABAQUS finite element software is used to describe its
mechanical behavior. The mechanical parameters of each microscopic component used in thispaper are shown in Table 2, and the values of each parameter are mainly taken from literature(Jayasuriya 2020).

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Table 2. Material parameters
Expansion
Group
Angle
BA(Mathew)30
NA(Jayasuriya)
30
FF-MC
30
ITZ(Jayasuriya)
15

Eccentricity

fc0/fb0

k

0.1
0.1
0.1
0.1

1.16
1.16
1.16
1.16

0.6667
0.6667
0.6667
0.6667Modulus ofelasticity
16000
70000
28960
26000

Poisson's
ratio
0.2
0.16
0.22
0.22

3.3. Numerical simulation result
Fig. 3. shows the variation law of FFF-BAC compressive strength with different BA content withsection size. As mentioned above, the compressive strength decreases with the increase of brickaggregate content and decreases with the increase of specimen size. The relationship between thecompressive strength of FFF-BAC and the specimen size was linearly fitted. The slope of thelinear fitting line is K, which is used to represent how fast the compressive strength of thereclaimed concrete changes with the increase of the specimen size. It can be seen that with theincrease of BA content, the K of FFF-BAC gradually increases, indicating that the size effect ofrecycled concrete is inhibited with the increase of BA content, which corresponds to the above
description.
Compressive strength (MPa)Compressive strength (MPa)45
K=-0.06515
40
35

R2=0.9597
30
25

Compressive strength(MPa)

45

50

40
K=-0.04932
35

30

R2=0.9631

25

20

20
100

200

300

Model size (mm)（a）BAC-0

400

100

200

300

Model size (mm)（b）BAC-50

400

32
30

K=-0.0246

28
26

R2=0.958

24
22
20
18
100

200

300

400

Model size(mm)（C）BAC-100

Fig. 3. FFF-BAC stress-strain curves with different BA substitution rates and different crosssection sizes
4. Conclusion
With the increase of BA content, the size effect coefficient increases, the slope K of the linear
fitting line between compressive strength and cross-section size gradually increases, and the size
effect of FFF-BAC is suppressed.
5. References
Sun X, Gao Z, Cao P 2019, Mechanical properties tests and multiscale numerical simulations forbasalt fiber reinforced concrete, Construction and building materials, 202: 58-72.
Zhang Li, Yu Zhenpeng, Shen Li, Sun Xue, Xie Xinghua 2019, Study on size effect ofcompressive strength of recycled concrete with different replacement rates, New Building
Materials,46(03):18-21+87.

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Su Jie, Zhu Jun, Shi Caijun, Fang Zhi 2022, Size effect on compressive strength of reclaimed
concrete strengthened by silica fume, Journal of Hunan University (Natural Science
Edition),49(05):174-180.
Zhou J, Pan J, Leung C K Y 2015, Mechanical behavior of fiber-reinforced engineered
cementitious composites in uniaxial compression, Journal of materials in civil engineering,27(1): 04014111.
Li Hedong 2009, Experimental study on Ultra-high toughness cement-based Composites , Dalian
University of Technology. (in Chinese)
Chen Yadi 2019, Study on Fiber pull-out Test and macroscopic Mechanical Properties of fiber
Reinforced cement-based Materials, Hefei University of Technology.
Mathew A A, Saibabu S, Mohan V 2021, Analysis of Strains in Brick Masonry Prism Using
ABAQUS, Proceedings of SECON 2020: Structural Engineering and Construction Management
Springer International Publishing, 883-897.
Jayasuriya A, Adams M P, Bandelt M J 2020, Generation and numerical analysis of random
aggregate structures in recycled concrete aggregate systems, Journal of Materials in Civil
Engineering, 32(4): 04020044.
Raj A, Borsaikia A C, Dixit U S 2020. Finite Element Modeling of Autoclave Aerated Concrete(AAC) Masonry for Estimation of Strength, Manufacturing Engineering: Select Proceedings of
CPIE 2019. Springer Singapore, 511-524.

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Finite Element Analysis Model of Prestressed Concrete Girder throughNonlinear Material Properties Correction of Strand
Do-Yeon Kim1, Young-Ji Park1, Jin-Woong Choi2, Young-Jin Kim3,
Il-Young Jang1, and Seong-Kyum Kim1*
1: Kumoh National Institute of Technology, Gumi, Korea; email: skim@kumoh.ac.kr: Korea Expressway Corporation Research Institute, Hwaseong, Korea; email: jchoi@ex.co.kr
3: Korea Concrete Institute Research Center, Seoul, Korea; email: kimcrete@naver.com
2*: corresponding author

Keywords: FEA; PS strand; PSC girder; Nonlinear material correction; Optimization
Abstract: This paper proposes a helical strand modeling for finite element analysis ofprestressed concrete structures, which is widely applied worldwide. Structural analysis using
nonlinear finite element analysis is being used to evaluate the performance of structures andpredict the behavior instead of structural experiments that require a lot of time and effort. For
this purpose, detailed modeling of all member elements is being built, but the helical strand, oneof the important members of the prestressed concrete structure, is difficult to construct a precise
model because of the structural shape and material specificity such as post-tension compressive
force within the elastic range. Therefore, in this study, the stress strain curve results from thetensile test results of prestressed strands used for prestressed concrete girder and tensile analysis
through Abaqus were used to correct the stress strain curve for material properties containing
non-linear behavior. The derived finite element model can be used as a model that shows similar
mechanical properties in a shape like a round steel bar, out of the complex helical strand
shape, and contributes to obtaining more efficient and accurate analysis results. In this process,the evolution algorithm used for optimization classifies objects that can approximate theobjective function, determines the reaction force according to the displacement through repeatedcalculations through re-analysis 50 times, and corrects it in the round steel bar model to obtain
the results of actual prestressed strands. Stress and properties of round steel bar model wereallowed to be applied. It is expected that these results can be utilized as a model that can reflectmaterial specificity while simplifying the model for prestressed structural analysis.
1. Introduction
PS strands are twisted strands of cold-worked steel wires after patenting and then bluing. PS
strands have high strength because they are heat-treated to form a sorbite-like pearlite structure,
which increases their tensile strength. However, Abaqus cannot implement the material
properties of twisted steel wire after cold working because it cannot simulate the chemicalcomposition transformation that occurs during heat treatment. If the material properties of the PS
strand are used in the material properties of the steel bar, the behavioral changes due todifferences in shape will not be reflected in the analysis. Therefore, the performance of the PS

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strand can be obtained by applying the corrected material model derived through nonlinear
material property optimization, and more accurate analysis results can be obtained when applied
to prestressed concrete girders. Additionally, Modeling the strand as steel bar not only reduces
penetration problems between elements, but also ensures convergence in analysis, improves
analysis efficiency, and makes modeling more convenient.
2. Analysis Preparation
In this paper, a finite element analysis of the strand and steel bar was performed using the
Abaqus program. The strand was composed of 7 wires, where a center wire was enclosed tightly
by 6 helical wires, resulting in a total diameter of 27 mm. The steel bar was also modeled with a
27 mm diameter, and the length of both the strand and steel bar was 2000 mm. The elastic
properties are shown in Table 1, and the plastic properties were set by referring to a previous
study (Chueachom, K. and Sirimontree, S. 2015).
Table 1. Material properties of strand
Parameter

Value

Density, kg/m3

7,850

Young’s Modulus, Pa

2E+11

Poission’s ratio

0.3
Fig. 1. Cross section of PS strand model

Due to its complex geometry, strand is more difficult to model and analyze using numericalmethods than normal steel bar. The strand is a shape in which several steel bars are twisted
together, and as it has many curved surfaces in contact, penetration between elements occurs inboth horizontal and vertical directions, even if the aspect ratio quality of the mesh is fine. (LinLu et al. 2022) Therefore, we used a 10-node modified quadratic tetrahedral mesh for the strand
model to effectively divide the model, and an 8-node linear hexahedral mesh for the steel bar
model. The interaction between each wire in the strand was defined by the General Contact in
Abaqus, and the contact properties were set as hard contact with a coefficient of friction of 0.2.
To apply the load history over time and derive a result over a time increment, the analysis was
conducted through the Dynamic/Explicit solver. At both ends of the strand and steel bar, 0.1 m
were tied by coupling to set a reference point, with one side fixed and the other side tensioned.
The tensile speed of the analysis was applied at a constant speed of 0.5 m/s.
3. Material Properties Correction
Nonlinear material property correction was performed using the optimization program Isight tocorrect the performance of the steel bar model by matching it to that of the prestressed concrete
girder strand. The correction components were set using Abaqus and Data Matching. In Data
Matching, the reaction force-displacement graph of the reference value strand was set as a target,
and the objective was to minimize the sum of the absolute area differences with the simulation
value of the steel bar. The technique used for correction in Isight was the Evolutionary374

Bridge Engineering Institute Conference 2023 (BEI-2023)
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Optimization Algorithm (Evol). The design variables were the Young's modulus and plastic
properties of the strand.
4. Results and Discussion
In Fig. 2(a), the strand and steel bar were initially stretched a total of 0.06 m at a speed of 0.5
m/s. However, it was determined that the model had fractured at the tensile point of 0.02 m after
reaching the ultimate strength and necking. Furthermore, the elements were abnormally distorted
after the fracture occurred. In a real experiment, if a load is applied after necking, the specimen
would be fractured. However, in finite element analysis, the elements can withstand the load
after necking without fracture. Since the value after fracture is meaningless in material property
correction and does not need to be optimized, the analysis and optimization were conducted up to
0.02 m to avoid considering the conditions after fracture. The results are shown in Fig. 2(b).

(a)

(b)

Fig. 2. Behavior of Reaction force-Displacement: (a) 0.06 m tensile behavior of 7-wires strand,
D27 steel bar; (b) 0.02 m tensile behavior of 7-wires strand, D27 steel bar and Material
correction No.45
In Fig. 2(b), the ultimate strength of the steel bar was 1,350 kN, which showed a 60%
difference compared to the ultimate strength of the strand of 800 kN. However, after material
property correction, the ultimate strength of the corrected steel bar was 700 kN, which was only
12.5% different from the target strand, and the trend of the reaction force-displacement graphbecame similar. In Fig. 3(a), the stress distribution according to the tension of the strand, steel
bar, and corrected steel bar was confirmed. In all three models, the maximum stress occurred at
the fixed end, and the corrected steel bar had a maximum stress that was 35% lower than thestrand. It was observed that the strand had maximum stress occurring at the fixed end, while thesteel bar had uniformly distributed maximum stress at all elements except the fixed end.
However, as a result of optimizing material properties, the stress distribution of the steel bar
tended to concentrate at the fixed end, similar to the strand. Out of the total 50 cases, minimized
objective result was derived in the No. 45. As shown in Fig. 3(b), the tendency of the plastic
curve of the corrected results was very different from the375

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reference. In addition to the plastic area of material properties, Young’s Modulus set withvariable decreased by 50% from 2e+11 to 1.08e+11 as a result of optimization. The reason forthe significant difference in material properties and trends from previous results is thought to be
due to the large difference in reaction force-displacement curves between the strand and steel
bar.

(a)

(b)

Fig. 3. (a) Stress distribution contour by tensile of the strand, steel, and corrected steel bar; (b)Comparison of plastic strain-yield stress curve between reference(Chueachom, K. andSirimontree, S. 2015) and corrected by Isight
5. Conclusion
In this study, the performance of PS strand and steel bar of the same diameter was analyzed
through finite element analysis, and a corrected material model for the steel bar was developed
through nonlinear material property optimization. It was determined that the Young's modulus
should be reduced and the trend change in the yield stress-plastic strain curve needed to beobtained for the performance of the PS strand of the steel bar. Therefore, the shape of the PS
strand can be simplified and by increasing the number of optimization cases, a more accurate
optimal material model could be derived.
6. Acknowledgment
This work was supported by the National Research Foundation of Korea(NRF) grant funded by the
Korea government(MIST) (No.2021R1C1C1013130)
7. References
Chueachom, K. and Sirimontree, S. 2015. “Prediction of Flexural and Degradation Behaviors ofFull Scale Prestessed Concrete Girder by Finite Element Method”, American Transactions on
Engineering and Applied Sciences, 4(3), 201-212.
Lin Lu et al. 2022. “The Effect of Lay Length on the Antipull Performance of Steel Cords andthe Spinning Loss Based on the FEM”, Advances in Materials Science and Engineering, 1-8.

376

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Corrosion and Durability

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Visualization of the Effect of Steel Corrosion Cracks
on Shear Deformation inside Mortar Using X-ray CT
Takayuki Fumoto1*, Koshiro Maekawa2, Shintaro Kikuchi3, and Yuta Yamada4
1: Kindai University, Higashi-Osaka, JAPAN; email: fumoto@civileng.kindai.ac.jp: Takada kiko Co.,LTD, Osaka, JAPAN; email: k_maekawa@takadakiko.co.jp
3: Kindai University, Higashi-Osaka, JAPAN; email: 2333350511j@kindai.ac.jp
4: Nihon University, Tokyo JAPAN; email: yamada.yuuta@nihon-u.ac.jp
2*: corresponding author

Keywords: Corrosion, Crack, Shear deformation, X-ray CT, Digital Volume, Correlation
Abstract: To evaluate the member capacity after the corrosion of steel bars, it is necessary toestimate the stress transfer mechanism based on the cracks inside the concrete induced by the
corrosion of the steel bars. However, the relationship between the spatial distribution of cracks
inside the concrete and the mechanism of stress transfer at those locations remains unclear. To
elucidate the stress transfer mechanism, it is necessary to visualize the crack locations inside thespecimen and to measure deformation under loading at those locations. Therefore, the purpose ofthis study is twofold: to demonstrate the effectiveness of X-ray CT images for visualizing theshear deformation near the corroding rebar and the cracks around it and to discuss the changes inthe stress transfer mechanism. The X-ray CT apparatus used in this study, which combines a load
testing apparatus and an X-ray CT unit, was developed by the authors. In this study, a mortar
specimen with three round steel bars placed within was prepared for shear testing. The internalcracks were observed from X-ray CT images of the specimens that had been subjected toelectrical corrosion. The specimens with and without electrolytic corrosion were then subjected
to shear tests, and three-dimensional deformation was measured by a digital volume correlation
method using X-ray CT images under different loads. After the round steel bars inside the mortar
were subjected to electric corrosion, the cross-sectional images of the specimens revealed the
reduction of the sectional area of the round steel bars and the formation of cracks around theround steel bars due to steel corrosion steel were observed from the cross-sectional images of thespecimens. In the absence of corrosion, stress transfer in the shear plane was continuous due toadhesion. In contrast, corrosion cracking in the shear plane changed the stress transfer
mechanism from adhesion to friction at relatively low loads.
1. Introduction
When assessing the performance of existing reinforced concrete structural members withcorroded reinforcement, it is essential to accurately estimate the stress transfer mechanisms in theconcrete. This requires information, such as surface crack widths and reinforcement distribution.
The stress transfer mechanisms, such as the reduction in adhesion stress of the reinforcement andthe decrease in apparent shear stress at the cracked surface, must be estimated, for example, fromsurface crack width and reinforcement arrangement. It is becoming increasingly clear that therelationship between the width of surface cracks and the amount of corrosion of the378

Bridge Engineering Institute Conference 2023 (BEI-2023)
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reinforcement is influenced by factors such as cover thickness. However, there are still many
unknowns, such as the apparent stress transfer mechanism due to corrosion cracking.
This study aims to demonstrate the effectiveness of an X-ray CT system that can capture images
under loaded conditions as a method for visualizing the relationship between corrosion cracking
patterns in the specimen and changes in the stress transfer mechanism in the vicinity of thereinforcement. To achieve this, we attempted to measure the shear deformation near the roundsteel bars with and without corrosion in the shear plane and discussed some aspects of thechanges in the stress transfer mechanism.
2. Specimen and Test Method
Table 1 shows the proportions of the mortar mixture, which consisted of tap water (W), earlystrength Portland cement (C), crushed sand (S1), and lime sand (S2). To improve accuracy inimage measurement, 30% of the fine aggregate unit volume was replaced with copper slag fineaggregate (Cu). The dimensions of the specimen, shown in Fig. 1, were 70 mm long, 70 mm
wide, and 210 mm high. Two horizontal slits with a thickness of 3 mm and a depth of 35 mm
created a 70 mm high, 70 mm wide shear plane (red box in Fig. 1) at the center of the height bycompressive loading. Three round steel rods, denoted as Rod A, B, and C, were placed in theshear plane. Seven specimens were created, and after demolding the following day, were cured
in water at a temperature of 20 °C for 14 days. Four specimens underwent an electrolytic
corrosion test, in which the right side of the specimen (Fig. 1(a)) was submerged in salt water toa height of 2 cm for 30 minutes and then energized for 18 hours at a voltage of 70 V and acurrent of 0.41 A. The average corrosion rate after energization was 9.68%. The specimens werethen cured in a room at a temperature of 20 °C and a relative humidity of 60% for 24 hours tostabilize the internal water content before subsequent tests. During the electrolytic corrosion test,
the three non-corroded specimens were also immersed in tap water and adjusted to the same
moisture conditions in a room under constant temperature and humidity.
Static loading tests were conducted on two specimens each without and with corrosion (Fig.
1(b)). The deformation of the specimen in the
Table 1. Mixture proportion of mortar
height direction was measured from the average
Unit content (kg/m3)
value of two displacement transducers as the W/C S/C
W
C
S1
S2 Cu
displacement of the shear plane. Characteristic (％)
points in the relationship between displacement 68
2.4 297 438 622 430 595
Conducting wire

L oading
place

Mortar35

placement

210
70

plane

Shearing plane( U nit:mm )A
CB RodL oading
place70(a) Specimen details

70

Slit
(Width = 3) 35
Steel rod (φ 5)70(A=2mm2)

70(b) Static loading test(c) X-ray tomography with loading
Fig. 1. Specimen and testing images

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and load during the static loading tests were identified as the target loads for X-ray computed
tomography (CT) imaging.
Next, one specimen each without and with corrosion underwent internal imaging using an
improved X-ray CT system owned by Kindai University under the target loads. The X-ray scansystem was located on the inward side of the loading apparatus with a capacity of 300 kN. The
imaging conditions were as follows: tube voltage 220 kV, tube current 100 µA, frame rate 3.0
frames/s, number of projections 2000, copper plate 2 mm, and resolution 0.0615 mm per side.
The X-ray CT images at 3 kN and other loads were used to measure internal deformation using
Digital Volume Correlation (DVC).

30

30

25

25

20
Case of Static loading test
Blue: Without corrosion
Red: With corrosion
Solid line: No.1
Dotted line: No.2

15
10
5
0 0

Load (kN)

Load (kN)

3. Result and Discussion
Fig. 2(a) shows the load-displacement relationship in the static loading test. No significant
difference was observed in the failure load between the corroded and uncorroded specimens. In
the case without corrosion, the load and displacement were proportional, and the brittle shear
failure occurred at a displacement of 0.4 mm. In the case with corrosion, the initial behavior was
similar to that without corrosion, but the displacement increased rapidly from 10 kN, and thespecimen failed at a displacement of 1.3 mm. The stress transfer mechanism may have beendifferent from that of the corroded steel bars, starting at about 10 kN. Therefore, the scanningtarget loads were 3 kN (initial value), 10 kN, 16 kN, 21 kN, and just before failure.
Fig. 2(b) shows the load-displacement relationship during X-ray CT imaging. The arrows in the
figure indicate the time intervals during which the loading plate position was held during the 11minute imaging period. However, due to the stress relaxation caused by fixing the loading plate
during the scan, the displacement of both specimens increased during re-loading after the scan,
as doing the cyclic loading. As shown in Fig. 2(a), the displacement of the specimen withcorrosion was greater than that of the specimen without corrosion.
Fig. 3 shows the horizontal cross-section of each specimen before and after 18 hours ofenergization, in which the round steel bar is displayed in white, the copper slag fine aggregate inwhite-gray, the mortar in gray, and the voids in dark gray. After the energization, the crosssection of the round steel bar was partially defective due to corrosion, and a yellow dotted line
indicated a shadow that appeared to be a crack on the right side. The amount of corrosion of eachsteel bar (Rod A, Rod B, Rod C) was almost the same.
Fig. 4 shows the displacement distribution in the loading direction inside the specimen. The
upper panel shows the horizontal cross-section at the center of the shear plane height (the upper
panel is the driven-in surface; the lower panel is the shear plane). The left half of the image isalmost fixed, and the right half moves upward (displacement is negative) due to loading from the

0.5

1
1.5
Displacement (mm)

20
15
10

Case of X ray CT
Blue: Without corrosion
Red: With corrosion

5
2

0

0

0.5

1
1.5
Displacement (mm)(a) Without X-ray CT scanning(b) With X-ray CT scanning
Fig. 2. Relationship between load and displacement3802Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

bottom of Fig. 1(a). In the case of no corrosion,
displacement increased continuously from left toright in both horizontal and vertical sections justbefore failure as in the case of 10 kN loading. In
other words, it is considered that stress wastransmitted on the shear plane until just before
failure. On the other hand, in the case of thecorroded specimen, the deformation became
discontinuous near the cracked area in Fig. 3 when (a) Before energization (b) After energization for 18h
a 10 kN load was applied, and the displacement Fig. 3. Horizontal cross-section image at thecenter of the height of the specimen
difference became continually larger till justbefore failure. In other words, The X-ray CT images confirmed that the change in the loaddisplacement relationship due to rebar corrosion shown in Fig. 2 was caused by a change fromadhesion to friction in the stress transfer mechanism induced by the corroded cracks.
Without corrosion
10 kN
23kN

28kN

10 kN

Horizontal

Rod A

With corrosion
23kN

25kN

Scale
0

Rod B
Rod C

20mm

-1.85
(mm)

Rod B
Slit

VerticalFig. 4. Vertical displacement in the cross-section of specimen (The yellow dotted line is the bar)
4. Summary
1) The X-ray CT images before and after the corrosion of the rebar showed the possibility ofobserving the location of cross-sectional reduction due to rebar corrosion and the internal cracks
around the corroded rebar.
2) DVC using X-ray CT images during loading confirmed that the change in the loaddisplacement relationship due to rebar corrosion was caused by a change in the stress transfer
mechanism induced by the corrosion cracks.
In the future, changes in the stress transfer mechanism under different corrosion cracking
conditions will be investigated.
References
Fumoto, T. 2013. Development of A New Industrial X-ray CT System and Its Application toCompression Test of Polymer Concrete, Journal of Japan Society of Civil Engineers, 69(2), 182191. (In Japanese)
Qiao, D., Nakamura, H., Tran, K.K. Yamamoto, Y., Miura T. 2015. Experimental and analyticalevaluation of concrete cover spalling behavior due to local corrosion, Journal of StructuralEngineering, 61A, 707-714. (In Japanese)381

Bridge Engineering Institute Conference 2023 (BEI-2023)
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A Change in the Failure Mode and Decrease in the Strength of RC Slabs
Due to the Progress of Frost Damage
Hiroshi Hayashida 1
1

Civil Engineering Research Institute for Cold Region, Sapporo, Japan, hayashida@ceri.go.jp

Keywords: Bridge slabs; Frost Damage; Deterioration Depth; Strength; Horizontal Cracks
Abstract: In Japan, many bridges were constructed from the 1950s to the 1970s. For this reason,over the next 10 years, more than 60% of bridges will be over 50 years old. Due to this aging,damage to the bridge slabs that directly support the traffic load has become a serious problem. In
particular, bridge slabs in cold regions deteriorate rapidly even with a small amount of traffic,
and there are many reports of occurrence of potholes. Frost damage is the biggest factor of rapid
deterioration of bridge slabs in cold regions. The deterioration progresses in the depth direction
from the upper surface. In order to prevent occurrence of potholes, it is necessary to understandthe relationship between the depth of deterioration due to frost damage and the decrease instrength of slabs. However, the relationship has not been clarified. In this study, to clarify the
relationship between the progress of deterioration and the decrease in strength of slabs that
undergo frost damage, a static loading test was conducted. The parameter for the test slab
specimens was the deterioration depth. In a specimen with a deterioration depth of 1/4 of the slab
thickness, the strength did not decrease; however, in a specimen with a deterioration depth of 1/2of the slab thickness, the failure mode changed and the strength greatly decreased. The factors
for the change in the failure mode were considered to be (1) the horizontal cracks from freezethaw cycles were already present before loading; and (2) Diagonal tensile cracks occurredbecause the portion of the specimen corresponding to the web of the beam deteriorated and thetensile strength decreased.
1. Introduction
In this study, to obtain the relationship between the depth of deterioration due to frost damage
and slab strength, RC slab specimens whose parameter was the deterioration depth were created,
and an experimental examination was done on these specimens using a static loading test.
2. Experiment Outline
2.1. Overview of the specimens
The dimensions, and reinforcing bar arrangement of the slab specimens are shown in Fig. 1. For
the concrete, air-entraining agent was not used and the water cement ratio was set at 65%. The
mix proportion is shown in Table 1. Data on the material properties of the concrete and thereinforcing bars are shown in Table 2. Table 3 shows the designations of the specimens and theexperiment parameter. The experiment parameter is the deterioration depth from the uppersurface of the slab specimen.

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mm

mm

D16

D16

(b) direction of bridge axis(a) Perpendicular to bridge axis

Fig. 1. The shape, dimensions and reinforcing bar arrangement of the slab specimen
Table 1. Mix proportion
W/C Airs/a(%) (%) (%)

3

mix ratio (kg/m )
W

C

S

G

65 2.0 48.6 171 263 904 947

Table 2. Material properties

Table 3. Experiment parameter

Concrete

Rebar

Name

Deterioration depthComp strength

Yield strength

N
H/4

40mm

H/2

80mm

2

（ N/mm ）25.8

2（ N/mm ）362

2.2. Freeze-thaw test
In the freeze-thaw test, the depth of deterioration was controlled by the following two methods.
1) As shown in Fig. 2, thermal insulation was installed on the surfaces other than the top surface,and cooling and heating of the specimen was done only on the top surface of the specimen.
2) As shown in Fig. 3, freeze-thaw action was applied only to the set range of deterioration depth,
and the frozen condition was maintained in the range below the set deterioration depth.
More precisely, as shown in Fig. 3, a temperature sensor was installed at a depth of 40mm forspecimen H/4 and at a depth of 80mm for specimen H/2. The temperature of the specimen was
adjusted between -18 and 0° C at the depth where the temperature sensor was installed in eachspecimen. Under these temperature conditions, the specimens were subjected to freezing in air
and thawing in water 100 times for specimen H/4 and 300 times for specimen H/2.

sinsulation

2.3 Static loading test
The specimen was simply supported at two sides, and the span was 1040 mm. A 100mm ×100mm steel plate was used as the loading plate. Concentrated loading was done at one point at
the center of the specimen.

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3. Static Loading Test Results
Fig. 4 shows the load-displacement curve for each specimen. Cracks on the cut surface, which
was in the direction of the span, and those on the bottom surface of the specimen are shown in
Fig. 5. The following two types of cracks are seen on the cut surface.
1) Wide cracks that were considered to be fracture surfaces (thick lines)
2) Horizontal cracks caused by freeze-thaw action (thin lines)

Load (kN)

300

H/4

200

N

H/2

100
0

0

2

4
6
Displacement (mm)8Fig. 4. Load-displacement curves

Fig. 5. Loading cracks in the cut surface and on the lower surface of the specimen
3.1. Specimen N
Specimen N reached a maximum load of 268kN at a displacement of 5.6 mm (the black line in
Fig. 4). Immediately after the maximum load was reached, the displacement suddenly increased
and the load suddenly dropped due to the failure. At this moment, as shown in the lower subfigure of Fig. 5(a), cracks that form the border of the punching shear cone occurred on thebottom surface of the specimen. In addition, as shown in the upper sub-figure in Fig. 5(a),loading cracks, which connected with the cracks of the punching shear cone, were found on the
cut surface.

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3.2. Specimen H/4
Specimen H/4 reached a maximum load of 274kN at a displacement of 5.4mm (the blue line in
Fig. 4). The maximum load and stiffness of specimen H/4 were almost the same as those of
specimen N, and no drastic decrease like that seen for specimen N was observed. For specimen
H/4, the load decreased and the displacement increased due to failure immediately after the
maximum load was reached, which were similar to what occurred to specimen N, but the degreesof these changes were not as marked as those for specimen N. As shown in the lower sub-figureof Fig. 5(b), cracks that form the border of the punching shear cone, which occurred on thebottom surface of the specimen N, did not occur. This is consistent with the fact that the cracks
on the cut surface did not reach the lower surface, as shown in the upper sub-figure in Fig. 5(b).
3.2. Specimen H/2
The load-displacement curve for specimen H/2 (the red line in Fig. 4) was completely different
from those of specimens N and H/4. Specifically, the curve of specimen H/2 was as follows. As
shown in Fig. 4, from the initial stage of loading to a load of 140kN (the displacement was 1.8mm), the decrease in stiffness was negligible, and the load-displacement relationship showedalmost the same trend as those for specimens N and H/4. However, when the load was further
increased, the displacement suddenly increased and reached the maximum load of 172kN at adisplacement of 3.2 mm. The decrease in maximum load was greater and the displacement at
maximum load was smaller for specimen H/2 than for specimens N and H/4. As shown in Fig.
5(c), cracks that form the border of the punching shear cone did not occur on the bottom surfaceof the specimen, nor did the loading cracks on the cut surface reach the bottom surface.
4. The Reason for the Strength Decrease in Specimen H/2
It is considered that the strength decrease was greater for specimen H/2 than for specimen N due
to the change in the failure mode caused by the following factors: (1) the horizontal cracks fromfreeze-thaw cycles were already present before loading; and (2) Diagonal tensile cracks occurredbecause the portion of the specimen corresponding to the web of the beam deteriorated and thetensile strength decreased. Because the specimen had no shear reinforcement against the
diagonal tensile cracks and because the horizontal cracks occurred in the area without reinforcing
bars, both types of cracks grew rapidly in the specimen. Therefore, it is considered that thestrength decreased when the displacement suddenly increased with no considerable increase inloading, and failure occurred.
5. Conclusions
To clarify the relationship between the progress of deterioration and the decrease in strength of
RC slabs that underwent frost damage, an examination based on the static loading experiment
was conducted. Slab specimens in which the deterioration depth was the parameter were created
for this experiment. The knowledge obtained in this study is summarized as follows.
(1) In the slab specimen that was subjected to freeze-thaw by the cooling and heating of only theupper surface, horizontal cracks were generated throughout the upper part of the specimen.
(2) The strength of specimen H/4, whose deterioration depth was 1/4 of the slab thickness, wasalmost the same as that of the sound specimen N. The strength decrease was greater for specimen
H/2, whose deterioration depth was 1/2 of the slab thickness, than for sound specimen N.
(3) It is considered that the strength decrease was greater for specimen H/2 than for specimen N
due to the change in the failure mode caused by the following factors: (1) the horizontal cracks

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from freeze-thaw cycles were already present before loading; and (2) Diagonal tensile cracks
occurred because the portion of the specimen corresponding to the web of the beam deteriorated
and the tensile strength decreased.

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Experimental Performance of Lightweight Aggregate ConcreteDue to Freezing and Thawing
Sangwoo Kim1*, Sardorbek Rustamov2, and Jinsup Kim3*
1: Dept. of Civil Engineering, Gyeongsang National University, Jinju 52828, South Korea; email:kimsangwoo@gnu.ac.kr
2: Dept. of Civil Engineering, Gyeongsang National University, Jinju 52828, South Korea; email:sardorbek060894@gmail.com
3: Dept. of Civil Engineering, Gyeongsang National University, Jinju 52828, South Korea; email:jinsup.kim@gnu.ac.kr*: corresponding author

Keywords: Lightweight expanded clay aggregate, Compression strength, Flexural tensile
strength, Direct tensile strength
Abstract: Concrete structures are not restricted in dimensions and shape; they are inexpensive
and highly capable, making them the most economical material in modern architecture. As the
21st century entered, the demand for skyscrapers increased, and concrete structures became taller.
It is a desire to have long columns or beams with a large span of large structures, but there is aproblem with increasing the weight of concrete as higher strength and durability are required.
Due to these problems, research on the performance improvement through lightweightaggregates is being actively studied. However, there still needs to be more research on theperformance change of concrete made of lightweight aggregates due to aging. Therefore, in thisstudy, the results of experiments on concrete manufactured from expanded clay aggregates were
presented under freezing and thawing conditions. Freeze and thawing were performed using thefreeze and thawing in air method with reference to ASTM C666, and 150,300 cycles were
performed. Compressive strength, flexural tensile strength, and direct tensile strength wereexperimentally evaluated. Specimens that performed freeze-thawing and specimens that did not
perform freeze-thawing were experimented on the same day to compare numerical differences
and analyze the effect of freeze-thawing on lightweight aggregate concrete.
1. Introduction
Concrete is a synthetic material composed of cement, water, aggregates, hybrid materials, andproperty improvement agents, and is the most widely used construction material by humans
applied to various construction structures. Recently, science and technology are rapidly
developing, and technological innovation in concrete as social infrastructure is also actively
progressing. In the past, concrete only required physical performance, but as social demands
have recently diversified, concrete with high performance, multifunction, and high durability,
such as super concrete, has been developed and applied. In addition, as demand for high-risebuildings increases in the 21st century, we want to make long columns or beams with a large387

Bridge Engineering Institute Conference 2023 (BEI-2023)
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span, but it causes problems in increasing the weight of concrete because higher strength anddurability are required. Due to this problem, research on reducing the weight of concrete
through weight reduction of aggregates is also being actively conducted. However, there are still
few studies on the aging of concrete using lightweight aggregates. Therefore, in this study, theperformance change was measured by deteriorating due to the freezing and thawing of concrete
using lightweight aggregates. With reference to ASTMC 666, freeze-thawing was performed
using the air freeze-thawing method, and 0,150 and 300 cycles were performed. Compression
strength, flexural tensile strength, and direct tensile strength were experimentally evaluated. The
specimen that performed freeze-thawing and specimens that did not perform freeze-thawing
were experimented on the same day to compare numerical differences and analyze the effect offreeze-thawing on lightweight aggregate concrete.
2. Experimental Program
2.1. Materials
In this study, ordinary Portland cement (Type 1) with specific gravity 3.15 was used as a binder,in compliance with the requirement of Korean Standard KS L 52001. The physical properties of
the cement are shown in Table 1. Lightweight expanded clay aggregate(LEA) was produced in
firepower plant ash. The LEA was applied as coarse aggregate to manufacture the lightweightaggregate concrete. The physical properties of the LEA are indicated in Table 2. Constituents of
LEA are summarized in Table 3. Tap water with a temperature of 20 ± 2 C was used to provide
excellent workability. To reduce the amount of water in the concrete, a superplasticizer wasapplied.
Table 1. Physical property of cement
Setting time/min
Density
Fineness
3
2(g/cm )
(g/cm )
Initial
Final
3.15
3,400
230
390
Table 2. Physical property of LEA
Density
Bulk density
3(g/cm )
(kg/m3)1.75
250-450

Compressive strength (MPa)
3-day
7-day
28-day
12.5
15.0
32.5

Min. size of aggregate(mm)3Max. size of aggregate(mm)15Table 3. Constituents of LEA(unit: %)
siO2

Al2O3

Fe2O3

CaO

MgO

K2O

Na2O

TiO2

66.7

18.3

4.57

1.37

0.65

1.63

1.71

0.78

2.2. Mixing Procedure
To maintain good workability and improve the concrete strength in the first step, the lightweightaggregate was cleaned and mud was totally removed from the surface of the aggregate.
Furthermore, to increase the resistance to damage from freeze-thaw cycles, the aggregate wasentirely saturated with water for 24 h. It was concluded by Kucharczykova (2010) that theporous aggregate (pre-saturation) has an important impact on the final strength and durability of
LC when the aggregate size provides pores in the aggregate that can be filled with water in the388Bridge Engineering Institute Conference 2023 (BEI-2023)
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pre-saturation process. Afterward, the aggregate was removed from the water and dried in the
air condition (sunlight and wind) fully. The surface of the aggregate was allowed to drycompletely. The mixture table used was from Manu (2017). However, 15% of the cement was
replaced by fly ash. The water-to-cement ratio of the concrete mixture was almost 0.35. Table 4is the mixing table of lightweight aggregate concrete. All the procedures followed for making
and curing the concrete in the laboratory, were according to the requirements of the ASTM C
192/C 192M-06 (2006) standard.
Table 4. Mixing table of lightweight aggregate concrete (unit: kg/m3)
CementWater

Fly AshAggregate

Superplasticizer

Sand

578

205

102

696

4.1

326

2.3. Test Method
There are three kinds (Compression, Tension, Flexural) of tests to value the strength of thelightweight aggregate concrete. The cylindrical concrete specimens (100 × 200 mm) prepared
for the compressive strength tests conformed with the ASTM C39/C39M-18 (2018) standard.
After 28 days, the specimens were tested on a universal testing machine with a 2,000 kN
maximum load capacity. The flexural tensile strength of the concrete specimens was determinedusing 100 × 100 × 500 mm rectangular prismatic beams with center-point loading according to
ASTM C293-02 (2002). Experiments on the direct tensile strength of LC and FRLC wereconducted using dog-bone specimens with dimensions of 575 × 200 × 100 mm. The
experimental method for the freeze-thawing cycles involved two sets of specimens. In the firstset were specimens pretreated by freezing and thawing. This first set of specimens directly after
curing 28 days (0 freeze-thaw) cycle for all categories of concrete was placed into the chamber
and then were subjected for (150 and 300 freeze-thaw) cycles to temperatures ranging from 18℃ to + 4℃. The second set was pretreated only by air drying, and these specimens were
cured in the air for the same period used for the freezing and thawing cycle pretreatment. The
frozen-thawed(aged) and air-dried(not aged) specimens were tested on the same day according to
the selected number of freeze-thaw cycles, where 150 cycled frozen-thawed specimens wereequal to 80 curing days for air-dried specimens, 300 cycled specimens equal to 130 curing days,
including standard 28 curing days of frozen-thawed and air-dried specimens respectively. The
difference between the strength of the frozen-thawed and air-dried (pre-treatment) specimens
tested for an identical number of freeze-thaw cycles, 150 cycles (80 days) and 300 cycles (130days) were recorded and compared.
3. Result and Conclusion
The result is summarized in Table 5. Each value represents the average value for each case. The
results of specimens pretreated by freeze-thawed (aged) and air-dried (not aged) in thecompressive tests are shown in Fig. 1 (a). As a result of the experiment, the not aged specimen
was cured for 80 days and the strength increased. But the aged specimen did not show anincrease in compression strength due to freezing-thawing of 150 cycles. After 80 days, thecuring was over. The results of specimens pretreated by freeze-thawed(aged) and air-dried(notaged) in the flexural tensile tests are shown in Fig. 1 (b). As a result of the experiment, not theaged specimen was cured for 130 days and the strength increased. But the aged specimen showsa little decrease in flexural tensile strength due to freezing-thawing. The results of specimens

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pretreated by freeze-thawed(aged) and air-dried(not aged) in the direct tensile tests are shown in
Fig. 1 (b). As a result of the experiment, not the aged specimen was cured for 130 days and thestrength increased dramatically after 80days. But the aged specimen shows a 33% decrease inflexural tensile strength due to freezing-thawing. This paper shows the influence of freezing andthawing on the lightweight aggregate concrete which is made with Lightweight expanded clay
aggregate. As a result of the experiment, the compressive strength, flexural tensile strength, anddirect tensile strength of lightweight aggregate concrete all increased due to curing after 28 days.
However, exposure to freezing and thawing strength of concrete did not increase or rather
decreased. Therefore, research is needed to improve the resistance of lightweight aggregate
concrete.

(a)

(b)

(c)

Fig. 1. Comparison of each test: (a) compressive, (b) flexural, (c) direct tensile
Table 5. Test result of all specimen (MPa)Type of test

0cycle(28-day)Compressive strength
Flexural tensile strength
Direct tensile strength

36.0
3.9
0.9

150cycle(80-day)
Aged
Not aged
36.5
45.3
3.7
4.6
0.8
1.1

300cycle(130-day)Aged
Not aged
35.8
45.3
3.8
5.0
0.6
2.2

4. References
B. Kucharczyková, Z. Keršner, O. Pospíchal, P. Misák, and T. Vymazal 2010, Influence ofFreeze-thaw cycles on fracture parameters values of lightweight concrete, Procedia Eng., 2(1),959–966
Manu S. Nadesan, and P. Dinakar 2017, Mix design and properties of fly ash waste lightweightaggregates in structural lightweight concrete, Case Stud. Constr. Mater., 7, 336–347

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Rust Evaluation of Weathering Steel Bridges by Optical Spectra in the Visible
and Near-Infrared Regions
Ryuichi Inoue1*, Rina Hasuike2, Hirokazu Furuki3 and Toshihiko Aso4
1

: Dept. of Civil and Environmental Engineering, Yamaguchi University, Ube, Japan; email:b002wdv@yamaguchi-u.ac.jp
2: Dept. of Civil and Environmental Engineering, Yamaguchi University, Ube, Japan; email:hasuike@yamaguchi-u.ac.jp
3: Nippon Koei Co., Ltd, Tokyo, Japan; email: a5962@n-koei.co.jp
4: Dept. of Civil and Environmental Engineering, Yamaguchi University, Ube, Japan; email:aso@yamaguchi-u.ac.jp*: corresponding author

Keywords: weathering steel, rust, optical spectra, dimensional reduction, random forest
Abstract: Monitoring of the rusting condition is necessary for the maintenance of weathering
steel bridges. Some research has been conducted on rust condition monitoring methods that
apply machine learning to digital camera images. However, to evaluate such rust conditions fromdigital camera images is difficult due to rust has a variety of color tones. To solve this issue, thisstudy focuses on wavelengths other than that available under visible light and examines a rustevaluation method on weathering steel using optical spectra in visible light and near-infrared.
This study attempts to classify into two classes of rust: rust with a thickness of 200-400 µm,
which is considered protective rust, and rust with a thickness of 400-800 µm, which isconsidered non-protective rust. In this study, a hyperspectral camera (HSC) is used to measure
optical spectra of weathering steel surfaces in every 1 cm square. From the first derivatives ofthese optical spectra, four wavelengths were selected that are valid for the two-classclassification. The reflection intensities at these four wavelengths were dimensionally reduced by
KPCA to produce 2D data. A classifier was constructed by applying random forest to these 2D
data and tested on 128 data and the discrimination accuracy was 100 %.
1. Introduction
Weathering steel that has been used to construct bridges has a corrosion prevention property byforming protective rust layer. Thus, periodic monitoring of the rust condition is necessary tomaintain the weathering steel bridges. The criteria of rust condition evaluation (JSSC, 2009) aremainly based on appearance. Therefore, the evaluation is not always quantitative, and in somecases the evaluation is not easy. To solve these challenges, some research has been conducted onrust condition easy-monitoring methods that apply machine learning to digital camera images(Sugita, 2022; Hasuike et al., 2021). However, it is sometimes difficult to evaluate the rustcondition from digital camera images because rust has a variety of color tones.
In contrast, as an evaluation method which is not affected by color tone, near-infrared
wavelengths have been focused on. Previous studies shows that optical spectra in near-infrared

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are more effective than digital camera images in detecting rust on carbon steel, and there is apossibility to construct a rust condition evaluation classifier for carbon steel (Kerf et al., 2022; W.
G. Rowley, 2018). It is thought that optical spectra in near-infrared may also be effective in
evaluating rust on weathering steel, but there have been few such studies.
In this study, a rust condition evaluation method for weathering steel by measuring
optical spectra in visible light and near-infrared and constructing a supervised learning classifier
is investigated. This study attempts to classify into two classes of rust: rust with a thickness of
200-400 µm (Class A), which is considered protective rust, and rust with a thickness of 400-800µm (Class B), which is considered non-protective rust.
2. Spectrum Measurement Method
Figure 1 shows the HSC (Specim IQ, Spectral Imaging) which was used to measure optical
spectra. Wavelength resolution is 7 nm, and it can measure optical spectra of 204 spectral bands
in the wavelength range of 400-1000 nm. The wavelength of 400-770 nm is visible light and
770-1000 nm is near-infrared. A barium sulfate plate with no specific light absorption was usedas the reference plate for measuring optical spectra, and the reflected intensity was calculated by
dividing the radiance value of the specimen by the radiance value of the reference plate.
Figure 2 indicates the experimental scenery. To ensure the reproducibility of theexperiment, the optical spectra were measured in a room with 0 Lux illumination except for thelight source. A halogen light (500 W) with a wide range of wavelengths from visible light tonear-infrared was used as the light source. The HSC was placed 20-30 cm away from thespecimen so that the entire specimen could be measured.
Figure 3 presents an example of the corroded weathering steel specimen. Left one has 15
cm length, 7 cm width and right one has 7 cm length, 7 cm width. It is difficult to measure rustthickness at the edges of the specimens, and the 7 × 15 cm specimens has a through-hole.
Therefore, the measurement ranges of optical spectra were 6 × 11 cm and 6 × 6 cm, respectively.
In this study, this measurement area was divided into 1 cm squares that used as one data. The
rust thickness at the center of each data was measured and used as the correct data for supervised
learning. There were 263 data in Class A and 265 data in Class B.

Fig. 1. HSC

Fig. 2. Experimental scenery

Fig. 3. Example of the specimen

3. Measured Optical Spectra, Spectral Pre-processing, and Evaluation of Rust Conditions
3.1. Optical spectra of weathering steel surfaces
Figure 4 shows examples of optical spectra and their first derivatives. For both classes, theoptical spectral curves have minima at wavelengths of 516 and 896 nm, and maxima at
wavelengths of 796 nm. The first derivative values at wavelengths of 569 and 694 nm are larger
than that at the surrounding wavelengths. There are similar trends for all data, however, thereflection intensities at these wavelengths varied from data to data.
It is considered that the all wavelength data may have information that not need to rustcondition classification. Therefore, the wavelengths effective for the two-class classification

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0.30Class AClass B

0.20
0.10
0.00
400 500 600 700 800 900 1000
Wavelength [nm]

-0.01
-0.00
-0.01
500

Class A
580

Class B
700

Wavelength [nm]Fig. 4. Examples of optical spectra and its first derivative

800

Reflection intensity

0.40First derivative

Reflection intensity

were selected from the wavelengths shown in Fig. 4. Figure 5 shows box-and-whisker plots of
the reflection intensities at wavelengths of 516 and 694 nm. The quartile ranges of reflection
intensity at wavelength of 516 nm for classes A and B are indistinguishable. On the other hand,the quartile ranges of reflection intensity at wavelength of 694 nm for classes A and B are
distinguishable. Similar trends were observed at the other wavelengths (569, 796 and 896 nm)shown in Fig. 4. Therefore, in this study, the wavelengths 569, 694, 796 and 896 nm are assumed
to be valid wavelengths (characteristic wavelengths) for the two-class classification.
0.40
0.30

Class A

0.20

Class B

0.10
0.00

516

694

Wavelength [nm]Fig. 5. box plots of reflection intensities

3.2. Dimension reduction
For the optical spectra of nutrients in soil, Qi et al. reported that constructing classifier using
dimension-reduced optical spectra can prevent overfitting (Qi et al., 2018). This method may
also be useful in the classification of the rust condition of weathering steel. Therefore, this study
used kernel principal component analysis (KPCA), a dimensionality reduction method applicable
to nonlinear data to reduce the dimensionality of the measured optical spectra. Applying KPCA
to the reflection intensity at each wavelength data, and the extracted first and second principal
components are used to construct a classifier. To clarify the effectiveness of constructing aclassifier using different number of wavelengths, following three types of classifiers areconstructed and compared: classifier constructed using all wavelength data, and using nearinfrared wavelengths data, and using the characteristic wavelengths data.
3.3. Classifier construction method
Classifiers are constructed using Random Forest (RF), an ensemble learning of decision trees
often used as a supervised learning algorithm for the two-class classification. The training dataconsisted of 180 data each, the validation data consisted of 20 data each, and the remaining data
(Class A: 63 data, Class B: 65 data) were used as test data.
4. Results and Discussions
Table 1 shows the discrimination accuracy. When the classifier was constructed with dataobtained by applying KPCA to reflection intensities at characteristic wavelengths, discrimination
accuracy was the highest (100. %). And, when the classifier was constructed with data obtainedby applying KPCA to reflection intensities at all wavelengths, discrimination accuracy was thelowest (66.5 %).
Figure 6 shows the first and second principal components obtained by applying KPCA to
the reflection intensities at three different wavelengths. The smaller the number of wavelengths
in the optical spectra before dimensionality reduction, the less the overlaps in the data group foreach class and the more clearly each class of data can be separated. As a result, classification
accuracy is higher. Therefore, when constructing a classifier to evaluate the rust condition of
weathering steel by optical spectra in visible light and near-infrared, it is necessary to select393

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wavelengths effective for classifying from the measured optical spectra and use data obtained byapplying KPCA to their reflectance intensity.
Table. 1. Discrimination accuracy

characteristic wavelengths

Accuracy64.1 %

84.4 %

100.0 %

-1.00

-0.00

-1.00
-1.00

A B

0.00

1.00

KPCA Component 1(a) all wavelengths

-1.00

-0.00

-1.00
-1.00

A B

0.00

KPCA Component 2

near-infrared wavelengths

KPCA Component 2

all wavelengths

KPCA Component 2

input data

1.00

KPCA Component 1(b) near-infrared wavelengths

-1.50

-0.00

-1.50
-1.50

A B

0.00

1.50

KPCA Component 1(c) characteristic wavelengths

Fig. 6. Relationship between the first and second principal components

5. Conclusion
In this study, an evaluation method for the rust condition of weathering steel by constructing RFclassifier using HS data in visible and near-infrared is investigated. This study focuses on the
pre-processing for the HS data combining of optical spectral wavelength selection anddimensionality reduction by KPCA. As a result, the discrimination accuracy was 66.5 % when
all wavelength data were used, and 83.6 % when the near-infrared wavelength data was used. On
the other hand, the discrimination accuracy at four characteristic wavelengths data (569, 694, 796
and 896 nm) screened from the first derivative of the measured optical spectra was 100.0 %.
6. Acknowledgements
This work was supported by JST SPRING, Grant Number JPMJSP2111.
7. References
H. Sugita. 2022. Development of rust appearance evaluation model for weathering steel using
deep learning, Bulletin of Hachinohe National College of Technology, 56, 43-50. (In Japanese)
H. Qi et al. 2018. Evaluating calibration methods for predicting soil available nutrients using
hyperspectral VNIR data, Soil and Tillage Research, 175, 267-275.
T. D. Kerf et al. 2022. Identification of corrosion minerals using shortwave infrared
hyperspectral Imaging, Sensors, 22, 1.
R. Hasuike et al. 2018. Classification of corrosion deterioration on weathering steel based on
CNN, Artificial Intelligence and Data Science, 2, J2, 813-820. (In Japanese)
Japanese Society of Steel Construction (JSSC). 2009. Applicability Assessment and Corrosion
Prevention Maintenance of Weathering Steel Bridges, JSSC Technical Report, 86.
W. G. Rowley. 2018. Hyperspectral imaging for detection of corrosion on intermediate level
nuclear waste containers, University of Birmingham. (dissertation)

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Effect of Measures for Inhibiting Galvanic Corrosion
in an Aluminum Alloy Bridge and Guard Railings
Yoshito Itoh1* and Tatsuya Kawabata2
1
2: Nagoya Industrial Research Institute. Nagoya, Japan, email: itoh@civil.nagoya-u.ac.jp: Japan Aluminum Association, Tokyo, Japan, email: t-kawabata@alkyo.jp*: corresponding author

Keywords: bridge railing, aluminum alloy, galvanic corrosion, stainless steel, zinc flake coating
Abstract: This paper is based on several field surveys and deals with a high durability aluminumalloy plate girder bridge “Kinkei-Bashi” built 61 years ago in Hyogo Prefecture, Japan. It focuseson the effect of measures taken against galvanic corrosion between aluminum and steel parts ofthe bridge, but also, secondly, between aluminum alloy joint members in bridge guard railings andtheir stainless steel fastenings. On-site surveys were carried out to assess the degrees of galvanic
corrosion in the guard railings. In the light of the survey results, galvanic corrosion exposure tests
were then performed on railing part samples over a five-year period at the Miyakojima
Atmospheric Exposure Test Site in Okinawa. Two types of exposure conditions, direct and undereave exposure, were tested for, using 250 specimens of 6000 series aluminum alloy joints.
Stainless steel fasteners with a zinc flake coating were found to be the best means for inhibiting
galvanic corrosion in aluminum alloy members in this kind of highly corrosive environment.
1. Introduction
Aluminum alloy is a lightweight material combining high durability with esthetic grace. Civil
engineers have used it since the 1980s in structures such as bridge railings, laterally attachedwalkways, high-water / tsunami barriers, land locks, and lighting masts. Recently, uses have also
been found for it in bridge inspection paths and work scaffolds, and the Japan Society of Civil
Engineers (JSCE) is preparing design and manufacturing guidelines for its use as a main material
in further types of structures such as emergency relief bridges. This paper opens with results froma survey of the present state of durability of the Kinkei Bridge (Japanese: “Kinkei-Bashi”), an
aluminum alloy plate girder bridge constructed 61 years ago, and then focuses in on corrosion tests
performed on samples of the fasteners used to connect 6000 series aluminum alloy joints to bridge
guard railings. The object is to investigate the effects of these fastener choices on durability and toverify their suitability as resources for inhibiting corrosion.
2. The Kinkei Aluminum Alloy Plate Girder Bridge Built in 1961
2.1. A general overview and on site survey
The Kinkei Bridge is the only aluminum alloy road bridge presently existing in Japan. Its visualand schematic appearance are shown in the photographs in Fig. 1 and the drawings in Fig. 2. Asseen in Fig. 2, it is a plate girder bridge of composite structure for live loads composed of fourmain girders, each with a span of 20.6m and a width of between 8.1m and 7.0m (the road has a395

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slope). It carries the “Ashiya Driveway”, a 10km stretch of toll road linking the two cities of Ashiya
and Arima Onsen in Hyogo Prefecture. The bridge was completed in June 1961 and on theoccasion of the 50th anniversary of this, the Durability Subcommittee of the Japan Aluminum
Association (Chair: Yoshito Itoh) discussed a proposal to nominate it for the 2017 JSCE
Engineering Heritage Award. The nomination was successful and the bridge was selected forinscription.
Corrosion-resistant rolled aluminum alloys plates (former JIS H 4104 A2P7-F, current
equivalent A5083) were used in the girders, and members were connected by MIG welding in thefactory (A54S-9/10, current A5654) and with rivets at the construction site. After the lapse of
more than 60 years, the girders are still in prime condition, a testimony to the durability ofaluminum alloy bridges. The main reason for choosing aluminum alloy as a material would havebeen the need to reduce the weight of the bridge in response to the soft clay ground in the Arima
area, but one can also imagine the urge the engineers must have felt to engage in this project.
Section A-A
Width 7120

Main Girder
Center Space

Left Bank

Right Bank

Fig. 1. The Kinkei Bridge

Fig.2. General drawings

Durability committee of the Japan Aluminum Association has been surveying the aluminum alloy
girders of the Kinkei Bridge periodically since 1995 (34 years after the bridge’s completion).
Latterly, these surveys have been at 5-year intervals, most recently in March, 2022 (61 years sincecompletion). In addition to a visual inspection of the bridge’s appearance, items surveyed included
amounts of salt deposition and depths of corrosion. As can be seen in Fig. 3, there are minor
incidences of pitting corrosion on both the outer and inner surfaces of the main girders, as well ason the bracings, lateral beams, and connecting joints. This is very shallow in depth, however,(0.6mm at most) and has hardly advanced in the past ten years. As main members, the aluminumalloy girders have remained in continuous service without any need of repair, an outstanding
display of endurance.

(a)Inner surface(b)Main girder end(c)Outer surface(d)Bottom flange

Fig. 3. Appearance of aluminum alloy main girder396

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2.2. Measures against corrosion from contact with steels, and their effectivenessAs reinforced concrete floor slabs, the Kinkei Bridge uses steel dowel bars with base plate and, as
shown in Fig. 4, thin plates of zinc, 1mm in thickness, are interposed between the dowel base andthe upper flange surface of the girder, to which they are attached with aluminum alloy rivets. As
zinc is a baser metal than aluminum alloy, a sacrificial corrosion effect is obtained from this. No
significant irregularities were evident even in 2022, 61 years after completion. The same insulation
methods were followed between the steel sole plate and the flange of the main girder thin zincplates of 1mm thickness were again interposed, as shown in Fig. 5. As fasteners, aluminum alloy
plate rivets and zinc plated bolts were used. Thanks to the sacrificial corrosion effect of the zincplates, corrosion is prevented on the lower flange of the aluminum alloy main girders, very severe
corrosion occurs on the cast iron bearings as iron is more noble than aluminum alloy. Indeed, after
the survey the committee recommended the replacement of this particular bearing. At any rate,
expedients such as interposed zinc plates can be seen as one way of guarding against galvanic
corrosion in places where different kinds of metal come into contact.

Zinc Thin Plate
Zinc Thin

Plate Plate
Sole

Fig. 4. Zinc plate between steel dubel and flange Fig. 5. Zinc plate between sole plate and flange
3. The Joints Durability Performance of Aluminum Alloy Guard Railings
Aluminum alloy parts can be joined in the factory by means of MIG or friction stir welding, buton construction sites, where welding is difficult, there is a need to use bolts. In an ordinary
environment where durability is not such a strict requirement, stainless steel bolts with no special
surface treatment are adequate, but in severe environments where durability is an issue, it isnecessary to use stainless bolts with a zinc flake coating (“Geomet🄬🄬” treatment). The need for thiscan be demonstrated experimentally, as argued below. Accelerated exposure tests was performed
using the 4 different nut / bolt / washer (etc.) assemblies shown in Fig. 6 (Mrema et al.2018; Itohet al. 2021). To obtain a quantitative measure of the effects of exposure, a long-term exposure test
(up to 5 years) was carried out at the Miyakojima Coastal Exposure Test Site in Okinawa
Prefecture. As shown in Fig. 7, two set-ups (direct exposure and eave exposure) were used for 14
types of variously shaped test pieces (250 specimens in total) representing joint members used in
guard railings.
After exposure periods equivalent to 1, 3 or 5 years, the bolts were disassembled andremoved and the alloy surface was minutely measured for corrosion using a laser depth meter. The
values recorded were the maximal depth of corrosion CDmax, the corroded surface area, and theamount (mass) of corrosion loss. Fig. 8 shows the outward appearance of one set of test pieces(aluminum plates) after 5 years of exposure. Fig. 8(b) shows the plates with the bolt assemblies
removed. Around the holes for bolts of type ‘A’ (stainless steel, non-coated), galvanic corrosion
can be seen resulting from the contact of the aluminum alloy with the steel.
The graph in Fig. 9 shows the progression in the depths of corrosion found with non-coated‘A’ type bolts after 1, 3, and 5 years of exposure. The maximal galvanic corrosion depth clearly
increases with the length of the exposure period. Moving to Fig. 10, however, which shows
corresponding results obtained with the use of zinc flake-coated bolts, no galvanic corrosion has
occurred even after the lapse of 5 years. As for ‘C’ type assemblies, in which only the washer is397

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painted for insulation leaving the bolt untreated, and ‘D’ type assemblies in which the bolt issurrounded by an insulating bush, both of these precautions are effective to a degree, but in bothcases galvanic corrosion continues to occur locally, due to paint peeling off in the case of type ‘C’or to deterioration in the bush material leading to a loosening of the insulation fit with type ‘D’.

Fig. 6. Specimen concept(a) Specimens with bolts

Fig. 7. Specimen set-up(b) Aluminum alloy plates removed bolts

Fig. 8. Specimen appearances after 5 years exposure

Fig. 9. Non-coated stainless steel bolt Fig. 10. Zinc flake coated stainless steel bolts
4. Conclusions
The measures taken against galvanic corrosion on the 61-year-old aluminum alloy plate girder
Kinkei Bridge were found to have been effective and to have resulted in a very high durability
performance. For connections between 6000 series aluminum alloy guardrails and railings,
stainless steel bolts with a zinc flake coating (“Geomet”🄬🄬) were tested and shown to give excellent
protection against galvanic corrosion by preventing contact between the two metals.
5. References
Mrema et al. 2018. Galvanic corrosion of aluminium alloy members of bridge guiderails under
severe atmospheric exposure conditions. Corrosion Engineering, Science and Technology, ISSN:
1478-422X (Print) 1743-2782 (Online), 54-2.

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Itoh et al. 2021. Durability of aluminum alloy members with fasteners using different surface
treatments against galvanic corrosion. Journal of Structural Engineering (JSCE), 67A, 431-442.

399

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Study of Water Absorption Properties in Actual Structure Concrete Based on
Pore Structure.
N’da Yacoub Bouadou1*, Peter Kuira Macharia2, Lai Lai Mon3, and Ryo Yoshida4
1: Nagoya Institute of Technology, Aichi, Japan; email: n.bouadou.525@stn.nitech.ac.jp: Nagoya Institute of Technology, Aichi, Japan; email: p.macharia.751@stn.nitech.ac.jp
3: Nagoya Institute of Technology, Aichi, Japan; email: l.lai.441@stn.nitech.ac.jp
4: Nagoya Institute of Technology, Aichi, Japan; email: yoshida.ryo@nitech.ac.jp
2*: corresponding author

Keywords: Durability; Existing concrete; Carbonation; Water penetration; Pore structure.
Abstract: Concrete is one of the most important materials in public works and building
construction, and because concrete is a living material, durability, and structural aging aregoverned by moisture content. However, it is difficult to determine the condition of reinforced
concrete because concrete is permanently evolving and exposed to various environments. This
paper provides an overview of design test methods in accordance with the JSCE and ASTM
Standard Specification for Concrete Structures, analyzes core samples taken from existing
concrete walls, and discusses future directions for new test procedures. Carbonation wasobserved in the collected cores and correlation between carbonation state and water infiltration
depth. This is thought to be an effect of coarsening of the pore structure due to carbonation.
Based on these results, the evaluation of water penetration in concrete should be based on thepore structure. Furthermore, the difference of the pore structure from the surface to the interior,of the real structures, needs to be considered in the moisture penetration prediction equation.
1. Introduction
Moisture penetration is the most significant factor in the deterioration of reinforced concrete. In
the durability verification of reinforced concrete in the Japanese Society of Civil Engineers, theverification method based on moisture penetration was established in 2017. However, thisverification method is tentative, and the prediction equation was developed through experiments
using homogeneous concrete produced in a laboratory. Therefore, this prediction equation cannot
accurately predict the depth of water penetration in concrete subjected to environmental
degradation as in real structures. This is because, in the concrete of real structures, there is adifference in the denseness (pore structure) between the surface concrete and the interior
concrete due to carbonation and drying, which is not considered in the prediction equation. The
objective of this study is to take into account the pore structure of the real structure due toenvironmental effects and to accurately predict the depth of water penetration in the real concrete
structure. To this end, water absorption tests (surface and immersion), carbonation tests, andPore volume rate were conducted on cylindrical specimens cored on existing road walls.
Specimens were analyzed in accordance with JSCE and ASTM standards. Based on the results ofthese tests, a new test method was devised by comparing the results with actual test methods andconsidering the distribution of densities that real concrete structures have.

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2. Outline of the Experiment2.1. Materials used and preparation of specimens
150mmx100mmø concrete cores were drilled from Tokai-dori line railway track barrier walls.
Visual inspection was carried done to sort out cores with open voids, cracks, and other visible
defects. The suitable cores were wrapped and stored in an open container at room temperaturefor 100 days. Using Rilem Tubes, an initial surface water absorption test was conducted on either
surface of each core to determine the effect of curing conditions and the effects of microcrack inthe concrete quality within 30min(Basheer et al.,1995) to evaluate the quality of concrete at thecover zone under natural dominant water suction. After the surface water absorption test, thecores were dried in an oven at 80°C until the mass change in 24hrs was less than 0.1%.
Table. 1. Average of sample dimensions and mass.
Specimen Properties
Diameter (mm)Thickness (mm)Water Temperature °C
Mass before sealing (grams)
Mass after sealing (grams)

JSCE
100
150
20
2655.822664.02

ASTM
100
50
20
827.08835.28W/C
60%

2.2. Experimental Methods
2.2.1. JSCE-G 582-2018 Test Standard
The following method was applied to determining water penetration rate coefficients of concrete
subjected to water in a short time. The concrete cores were dried in a controlled chamber at atemperature of 40 ± 2°C and relative humidity of 30 ± 5% until 24- hour mass change was lessthan 0.1% and cooled to room temperature by natural heat loss for of 1 hour and then stored in aclosed container for 24 hours. The weight of each dry core was measured and a 10 mm markfrom the bottom of the specimen was labelled and carefully placed 5 mm height. Tap water wasused as a test solution and the specimens were split at 48 hours after immersion and sprayed withan indicator. The water penetration depth was measured at 5 points using a caliper, and theaverage value was taken.
2.2.1. ASTM C1585-3 Test Standard
The test method used was ASTM C1585 – 03: “Measurement of Rate of Absorption of Water byHydraulic Cement Concrete”. This test method determines the rate of absorption (i.e. sorptivity)of water by hydraulic cement concrete by measuring the increase in the mass of a specimen
resulting from absorption of water as a function of time when only one surface of the specimen is
exposed to water. The disc concrete specimens of 100 ±6 mm diameter with a length of 50±3
mm were cut from the 150mm cores in order to have both internal and external disc concrete andvacuum saturated for 24 hrs, The specimen for 9 drilled cores were used were sealed withaluminum tape and later conditioned in an environmental chamber with a temperature of 50 ± 2 ̊
C and RH of 80 ± 3 % for 3 days. Each sample was then placed in a sealed container at 23 ± 2 ̊ C
for 15 days figure illustrative

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3. Experimental Results and Discussion
3.1. Relationship between JSCE and ASTM water absorption rate
In this study, the comparison between both internal and external (side exposed to carbonation) of
ASTM cores has been made with the JSCE cores in order to determine the water absorption rate
(sorptivity) of both the outer and inner concrete surfaces following different conditioning
methods [1]. Despite the fact that at the initial stage internal core of ASTM has a strong
correlation with JSCE external, we observed a strong correlation between JSCE methods and
ASTM methods Fig. 1.

Fig. 1. Relation between JSCE and ASTM water absorption rate
3.2. Relationship between water absorption rate and carbonation depth
After slicing the JSCE cores and after checking the water infiltration depth, carbonation wasdetermined in those cores using the phenolphthalein test method. Following the result of themesearment and as per the previous research, a correlation was observed between water
absorption and carbonatation state Fig.2 .The changes in water penetration resistance due tocarbonation were because of the modifications in the pore structures. It is considered that the
carbonation makes the concrete pore larger and consequently affects the water infiltration rate[2] .

Fig. 2. Relation between absorption rate and carbonation
3.3. Relationship between JSCE water absorption rate and core slice absorption rate.
From the results of the water penetration test, it is considered that the changes in water
penetration because of the porosity were calculated by Eq.1. The sample was immersed in
acetone for 24 hours with mass change recording after 1min, 5, 10, 15, 30, 60mins, the mass was
also recorded at an interval of one hour up to six hours.

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𝑤𝑤𝑎𝑎 = �(1)

𝑚𝑚1 − 𝑚𝑚2
� 𝑥𝑥100
𝑉𝑉

wa= Absorption rate (%)
m1 = mass in air,
m2 =absolute dry mass (g),v = volume of the specimen, in mm2

Fig. 3. Relation between absorption rate and carbonation
Based on the core slice study we observed a strong correlation between slice1-2 absorption rate
and JSCE absorption rate, which can tell this slice which has a high of 30mm represents theeffective water infiltration depth.
4. Conclusion
In this paper, the effect of the environment on real concrete structures has been investigated andit was confirmed that, despite the fact that ASTM and JSCE have different conditioning methods
they have a good capillarity suction rate correlation. Studying the cores slices gives relevant
information on the mechanism of water absorption in the existing concrete considering that realconcrete is heterogenous from the surface towards the interior due to the environmental effect.
Furthermore, pore structure can be considered as a durability indicator for further research onwater infiltration depth prediction.
5. Acknowledgments
We would also like to thank the laboratory colleagues for their help.
6. References[1] R. J. Gummer son, C. Hall and and W. D. Hoff, "Water Movement in Porous Building
materia;. ls --II. Hydraulic Suction and Absorptivity of brick and other Masonry Material," inWater Movement in Porous Building Vol.15, 1980, pp. 101 - 108.
[2] S. Maehara, M. Suzuki and K. Hayakawa, "Effect of Carbonation Depth on Short-Term
Water Penetration In Concrete," Tokyo Construction Technology Research Institute Bulletin
No. 46, Tokyo, 2018.

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Performance and Evaluation

Bridge Engineering Institute Conference 2023 (BEI-2023)
Rome, Italy, July 17-20, 2023

Evaluation of the National Bridge Inventory (NBI) to Predict Bridge
Deterioration in Missouri, USA
Glenn Washer1, John J. Myers2*, and Mohammed Hammed1: University of Missouri - Columbia, Columbia, Missouri, USA; email: washerg@umsystem.edu: Missouri University of Science and Technology, Rolla, Missouri, USA; email: jmyers@mst.edu*: corresponding author and plenary presentation
1
2

Keywords: NBI Inventory, bridge deterioration, bridge condition rating (CR), bridge service
life.
Abstract: In the state of Missouri, USA, The Missouri Department of Transportation (MoDOT)
is responsible for the inspection and maintenance of approximately 10,000 bridges and culverts on
the state roadway system. The condition of these structures is assessed biannually to monitor
deterioration that occurs as a result of environmental exposure and traffic loading. To effectively
manage these important assets, MoDOT launched a research effort to develop deterioration models
that would allow them to better project future preservation and rehabilitation activities and develop
data-driven asset management plans.
This study developed deterioration curves for different bridge components and culverts, identified
trends in deterioration patterns, and develop recommendations for cost-effective bridge types. To
conduct this analysis, records from the Federal Highway Administration’s (FHWA) National
Bridge Inventory (NBI) were obtained for the years of 1983 - 2019. These records document theinspection results from biannual inspection through condition ratings (CR) that describe thecondition of bridge components (deck, superstructure, and substructure) and culverts on a 0 - 9
numerical scale. This paper will discuss the key takeaways from the study in terms of influencing
factors, rate of deterioration once certain CR’s occur and future recommendations in terms of datacollection.
1. Introduction
Historically, bridge inspection in the US began shortly after the collapse of the Silver Bridge on
December 15, 1967. Following the collapse, requirements for the inspection and reporting ofbridge conditions to the FHWA were established in 1971, with the first round of inspection on
National Highway System bridges due in 1973. Inspections were completed on a component levelbasis that rated three primary components of a bridge (deck, superstructure, and substructure).
Additional data on the characteristics and location of the structure were also recorded. In
conventional component-level bridge inspection, bridge managers use inspection results to
identify maintenance needs from the safety perspective to help make decisions. A subjective CR
scale ranging from 0-9 is used to characterize the condition of bridge components relative to the asbuilt condition. Culverts receive a single CR to describe the general condition of the culvert. The
CRs are recorded on a 0-9 scale where 0 is a structure in “failed condition – out of service” and 9
is a structure in “excellent condition.” Structures in “good” condition have CRs of 7 or greater,
structures in “poor” condition have ratings of 4 or less. Structures with CRs of 5 or 6 are considered405

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in “fair” condition. The results of the inspection along with other data describing a structure aresubmitted annually by bridge owners to the FHWA. These data are stored in the NBI. Although
this data collection has been ongoing since the early 1970s, the available data from the FHWA
begins in 1983.
2. Statistical Analysis Methods
Improving the understanding of the deterioration and life expectancy of bridges allows fortransportation agencies to improve their planning processes and also better understand how certain
bridge types perform under various environmental and physical exposure conditions. In this work,
deterioration curves were developed based on Kaplan-Meir (K-M) method of survival analysis(Washer et al. 2014; Washer et al. 2018; Kaplan and Meier 1958). This methodology applies the
National Bridge Inventory (NBI) Condition Rating (CR) data collected over a time interval of 37
years to calculate the probability of likelihood of a structure to transition to the next lower CR.
These data were then analyzed and median service life estimates were provided for the deck,superstructure, and substructure of bridges formed from the materials of steel, reinforced concrete(RCC), and prestressed concrete (PSC). Similar curves and service life estimates were produced
for concrete culverts.
Cox regression analysis was used to identify and quantify trends in the deterioration of different
types of bridges MoDOT has built over the years. These data were analyzed to determine theinfluence of parameters such as the effect of salt application, environment, span length, traffic
volume, location, and age. For the purposes of this limited paper, the authors will focus onpresenting variables investigated, their general influence and data collection gaps.
3. Data for Analysis
Inspection data (records) for state-owned bridges and culverts in Missouri, USA were acquired fromtwo different sources within the FHWA. Records from the years of 1992–2019 were obtained fromthe Long-Term Bridge Performance Program (LTBP) web resource called InfoBridge (see Fig. 1).
In addition to these data, NBI records were acquired from 1983–1991 from the FHWA Office ofBridges and Structures. Data recorded in the NBI for the years of 1983–2019 for structures in
Missouri were affected by various changes in the numbering system used for state bridges. As aresult, data preparation included resolving numbering inconsistency in the records such that thehistorical performance of each bridge could be tracked over time. In total, 37 years of inspection
data were obtained for analysis. Fig 2. Shows the years of inspection data used for bridges in
Missouri, USA.
Greater than 50% of
structures have 36/37 years or
greater of inspection data.

Fig. 1. FHWA InfoBridgeTM Website

Fig. 2. Years of Inspection Data for MO Bridges

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The independent variables included the age of the structure, with consideration for renovation or
replacement. The structure kind, structure type, construction material (e.g., steel, reinforced
concrete, prestressed concrete, etc.) and construction type (e.g., simply supported, continuous,truss, girder, etc.). There are 56 different combinations of superstructure materials and designs,
including 51 different bridge combinations. In addition, structure length, span length, deck type,traffic volume (ADT and ADTT) were important parameters to examine. Other data such as
climate (e.g. freeze/thaw data by district/region in the state) and salt usage by county was collected
separately. Salt usage was converted to salt lanes miles in each district. The CR for decks,
superstructures, substructures, and culverts are the main dependent variables used to determine thereliability of structures in the MoDOT inventory.
4. Results and Discussion
Reliability analysis for CIP decks and service life plots were developed based upon the analysis
for all Missouri bridges in the database. Unlike some regions in the USA, superstructure decks inparticular are most suspectable to deterioration due to heavy deicing salt usage in the winter. Fig.
3 shows the reliability of CIP decks for CR 3-8. This can be used can be used to construct service
life estimates for CIP decks based on the percentage of the CIP decks still surviving at the medianvalue from the reliability plot. The median value is illustrated on the plot with a horizontal line at
50% reliability. Fig. 4. Shows the resulting service life for CIP decks based on medium reliability
in CR 3-8.

Fig. 3. K-M Curve for CIP Decks CR 3-8

Fig. 4. Service Life for CIP Decks in CR 3-8

Table 1. Table showing maximum likelihood estimate for CIP deck covariates
Var. no.

Variables in model

0
1
2
3
4
5
7
8
9
10
11
12

None
Age in TICR
Structure length (ft.)
Maximum span length (ft.)
Freeze/thaw (days/year)
Snow (days/year)
Salt (tons/lane miles)
ADT
ADTT
District
RegionSuperstructure type-2log 𝑳𝑳𝑳𝑳�

242,492.64
242,446.15
242,479.71
24,2435.75
242,455.34
242,463.27
242,419.31
242,492.25
242,492.41
242,398.57
242,442.91
242,228.75

407

Likelihood ratio

Result

46.49
13.99
56.89
37.30
29.36
73.33
0.39
0.23
94.07
49.73
263.89Null model
Significant
Significant
Significant
Significant
Significant
Significant
Not significant
Not significant
Significant
Significant
Significant

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Table 1 indicates the maximum likelihood estimate for CIP deck covariates. The most influential
factor, other than the superstructure type, which affected bridge deterioration in Missouri was theuse of salt. This may seem antidotal to most bridge engineers, but confirmation of this correlation
was an important finding. Fig. 5 and Fig. 6 show the Missouri district and relationship between
snow days and salt usage in Missouri.

Fig. 5. Missouri’s Seven Districts

Fig. 6. MO Salt versus Snow Days for 2001-2010

Naturally the type of superstructure had an important role in the NBI condition rating anddeterioration rate. Fig. 7 presents the service life
plot for different types of super structures. It may
be noted that Reinforced Concrete continuous(RCC) slab bridges (among the oldest in the system
network) followed by Prestressed Concrete
continuous (PSC) girder bridges exhibited thelongest service life in the database.
5. Concluding Remarks
The following concluding remarks are presented.
The type of superstructure resulted in different
service life expectations. The quantity of salt
application had greater impact on the deterioration of
structures than any other parameter studied. The
Fig. 7. Service life by Superstructure Typedata showed that ADT and ADTT did not have astatistically significant effect on the reliability of structures. An improved understanding of theservice life and major influencing factors in bridge life, will significantly assist the owner in systemselection, predicting structure life and planning. The following recommendation is suggested: amore precise tracking system to track the frequency and quantity of salt application is recommended
in each state in the US. This should also be integrated within the NBI data collection system.
6. References
Washer, G., et al., Proposed Guideline for Reliability-Based Bridge Inspection Practices. NCHRP
Report 782, 2014 (Washington, D.C.).
Washer, G., et al., Guidelines to Improve the Quality of Element-Level Bridge Inspection
Data. 2018, NCHRP: Washington, D.C.
Kaplan, E.L. and Meier, P., Nonparametric Estimation from Incomplete Observations. 1958,
53(282): p. 457-481.

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Assessment of Live Load Carrying Capacity of Prestressed Concrete Bridge: a Case Study
Shin-Tai Song1* and Yiching Lin2
1

: Associate professor, Department of Civil Engineering, National Chung-Hsing University,
Taichung, Taiwan, R.O.C.; email: ssong@nchu.edu.tw
2: Professor, Department of Civil Engineering, National Chung-Hsing University, Taichung,Taiwan, R.O.C.; email: yiching@dragon.nchu.edu.tw*: Corresponding author

Abstract: A prestressed concrete girder bridge is selected to illustrate the versatility of an approach
for assessing the live load carrying capacity of the bridge span. The prestressing force on a girder isestimated using a strain relief approach. The effective flexural rigidity of the girder is assessed byconducting a truck load test for the bridge span. Determining the flexural rigidity and prestressing
force will enable the construction of a finite element model of the bridge under its current condition.
The extent of the allowable stress increment of the bridge span is established by comparing the stresson the girders under their current condition with the stress limit for satisfactory performance. The
allowable stress increments are further used in bridge safety monitoring.
1. Introduction
The serviceability of prestressed concrete bridges is highly dependent on the stress level of theirtendons. The shrinkage and creep of concrete as well as the relaxation of the prestressing tendons
can cause prestress loss, resulting in the deterioration of the serviceability of a bridge. Additionally,
chemicals in the environment, overweight vehicles, and unexpected natural disasters may exacerbate
the deterioration of a bridge, resulting in unexpected prestress loss and the subsequent degradationof its serviceability. The actual level of prestress loss in the tendons cannot be estimated by simply
observing the damage on the girders or applying theoretical calculation. Rather, the prestress loss
must be determined through mechanical testing. The identification of the forces remaining on theprestressing tendons will allow for the assessment of the performance of a bridge. This study
demonstrates an approach for determining the prestressing force on tendons and evaluating the live
load carrying capacity of bridges with prestress loss. The primary consideration for utilizing theapproach is to comprehend the behavior of a bridge with limited instrumentation. Two sets of tests
are performed onsite to determine the effective flexural stiffness and prestressing force of the girdersof a bridge. Knowing the effective flexural rigidity and prestressing force will enable the constructionof an analytical model, which can simulate the load-deformation response of the bridge under its
current condition. By comparing the stress on the girders under the current condition with the stresslimit for the satisfactory performance of the prestressed concrete girders, the allowable stressincrement of the bridge span can be established. The allowable stress increments are further used inbridge safety monitoring.

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2. Bridge Model
A multi-span prestressed concrete I-girder bridge located on the West Coast Expressway in Taiwan
is selected to illustrate the versatility of the presented approach. Routine inspection of the bridgefound that several girders suffered from flexural cracks. As the bridge is located in a high-corrosionrisk region, the maintenance agency is concerned about the serviceability of the bridge and wants toassess its load carrying capacity. The bridge has 40 simply supported spans, with a span length of 40
m. Each span consists of eight prestressed concrete girders, which are connected by seven transverse
diaphragms. Each girder contains six prestressing tendons, which are made of twelve 12.7-mm
diameter strands. The effective prestressing force on a girder is designed to be PD = 7000 kN ,considering prestress loss. The ultimate tensile strength of a strand is f pu = 1870 MPa, and thecompressive strength of the concrete is f 'c = 35 MPa. The bridge span is modeled by a grillage
model, which idealizes the superstructure as a skeletal structure consisting of interconnecting
elements. The longitudinal elements simulate the response of the girders subjected to external loads,
and the transverse elements simulate the load distribution effect between the girders. The grillage
model of the bridge span is shown in Fig. 1. The superstructure of the bridge is simply supported,
allowing each span to exhibit independent behavior when subjected to external loads. The load
carrying capacity of each span can be assessed separately. For the purpose of this study, only one
span is selected to illustrate the assessment process.

Fig. 1. Grillage model of bridge span
3. Truck Load Test and Stress Relief Test
A truck load test and stress relief test are conducted onsite to assess the flexural rigidity andprestressing force of the girders, respectively. The truck load test is conducted by slowly moving
a 245 kN truck along girder 5 (G5) and briefly stopping at the midspan before proceeding to theadjacent span. Two strain gauges are installed at the midspan of G5, one mounted on the bottomof the girder and the other at 53 cm above the bottom of the girder. Fig. 2(a) shows a strain
increment of=εb 13.49 µε at the bottom of the girder when the truck was in position. Fig. 2(b)shows that the truck caused a strain of =ε w 7.63 µε at 53 cm above the bottom of the girder.
Following the assumption of a linear strain profile under flexure, the neutral axis of the girder
section is hn = 134 cm from the bottom. The analytical model of the bridge span shows that when
the span is subjected to a 245 kN load at the midspan of G5, the maximum bending moment of G5
is M mid = 325 kN-m . By determining the location of neutral axis hn , strain at the bottom of thegirder εb , and associated bending moment M mid , the effective flexural rigidity of the girder canbe calculated by410

Bridge Engineering Institute Conference 2023 (BEI-2023)
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EI e =

hn
M mid
εb

(1)

7
2
and is equal to EI=
e 3.23 × 10 kN-m . The effective flexural rigidity is 1.5 times larger than thetheoretical flexural rigidity estimated by the concrete strength f 'c and section geometry. The
prestressing force on the girder is assessed with a strain relief approach. The approach involves
cutting two slots into the web of G5, that is, one near the top flange and another near the bottom
flange. The compressive stress on the concrete will cause the edges of the slots to deform toward
one another. The deformation of the edges is measured to further calculate the stress on theconcrete (Jiang 2018). The top slot, which is 149 cm from the bottom of the girder, is measured ashaving a compressive stress of σtotal=
2622 kPa . The bottom slot is 82 cm from the bottom of thetsgirder and has a compressive stress of σtotal
2725 kPa . In addition to the prestressing force andbs =self-weight, the temperature gradient across the girder section will exert stress on the concrete.
This stress must be assessed to accurately evaluate the prestressing force on the girder. Figure 3
illustrates the temperature gradient measured during the strain relief test and the resulting stressinduced by this gradient. The temperature gradient causes a tensile stress of σtemp=
884 kPa onts
temp
the top slot and σbs =345 kPa on the bottom slot. By subtracting the stresses induced bytemperature gradient from the results of the strain relief approach, the stresses caused by theprestressing force and self-weight are obtained as σtsmech =3506 kPa , in compression, on the top
mechslot and σbs
3070 kPa , in compression, on the bottom slot. The stress on the bottom of the=girder is estimated to be σbmech =2530 kPa . Then, the prestressing force is determined to be
Pe = 5750 kN , which is 82% of the designed effective prestressing force.

14

6

13.49

Strain (µε)

10

11.83

2
-2 0

25

50
75
Time (s)

100

10
6

7.63

2
-2 0

(a)

25

7.63

50
75
Time (s)Fig. 2. Variation of the strain in truck load test
250

250

200

200

Location

Location

Strain (µε)

14

150

150

100

Compression
Top slot

100

50
0

Tension50

30 33 36 39 42
Temperature (°C)

Bottom slot

0-2000 -1000 0 1000 2000
Stress (kPa)
(b)

(a)

Fig. 3. (a) Temperature gradient measured in G5; (b) induced stress

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4. Live Load Carrying Capacity and Strain Increment for Bridge Safety Monitoring
7
2
The effective flexural rigidity of EI=
e 3.23 × 10 kN-m estimated from the truck load test andprestressing force of Pe = 5750 kN assessed by the strain relief test are further utilized to revise
the analytical model of the bridge under its current condition. The results of the stress relief test
also indicate that G5 is currently under a compressive stress of σbmech =2530 kPa at the bottom of
the girder when subjected to the self-weight of the bridge and prestressing force. For a prestressed
concrete bridge located in a high-corrosion-risk region, the bridge design specification (MOTC 2020)
indicates a tensile stress limit of ft a = 0.249 f 'c , which is 40% of the modulus of rupture ofconcrete and equal to ft a = 1470 kPa in the sample bridge. The difference between the compressive
stress currently applied to the girder and tensile stress limit provides an allowable stress incrementof 4000 kPa. The bridge was originally designed with a prestressing force of PD = 7000 kN . The
self-weigh of the bridge and prestressing force of PD = 7000 kN induce a compressive stress of
5160 kPa at the bottom of G5, and the allowable tensile stress increment is 6630 kPa. The
σbmech =comparison between the current allowable stress increment of 4000 kPa and stress increment of
6630 kPa in the original design shows that the live load carrying capacity of the bridge span is
reduced to 60%. The stress increment of 4000 kPa is equivalent to an allowable tensile strain
increment of =εta 97.1 µε at the midspan of G5. For the purpose of bridge safety monitoring, astrain gauge can be installed at the midspan of G5. A strain increment of =εta 97.1 µε can beapplied to verify whether the traffic load caused the stress level of the girder to exceed the limit
specified in the design code.
5. Conclusion
This study presents an approach for assessing the live load carrying capacity of a prestressed
concrete girder bridge and establishes the allowable strain increment for bridge safety monitoring.
A truck load test and strain relief test are performed onsite to determine the effective flexural stiffness
and prestressing force of a bridge girder. The results of the onsite tests will enable the constructionof a finite element model of the bridge under its current condition. The comparison between the
stress currently applied to the girder and stress limit specified in the design code can provide theallowable stress increment for the bridge span, which is further used to evaluate the live load carrying
capacity. For the bridge selected in this study to illustrate the versatility of the approach, the current
live load carrying capacity is 60% of the originally designed capacity.
6. Acknowledgements
This research was sponsored by the Maintenance Agency of District-2, Directorate General ofHighways, Taiwan, R.O.C. The financial support is gratefully acknowledged.
7. References
Jiang, Kuo-Ruei 2018. Numerical and experimental study of strain relief measurement method forprestressed concrete structures, Master Thesis, National Chung-Hsing University.
Ministry of Transportation and Communications (MOTC) 2020. Highway Bridge Design
Specifications, Taipei, Taiwan, R.O.C.

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Strategic Asset Management of Locally Owned Bridges
Mi G. Chorzepa1*
1*: The University of Georgia, Athens, Georgia, U.S.A.; email: chorzepa@uga.edu*: corresponding author
Keywords: locally owned; element; bridge; maintenance; priority
Abstract: In the United States, locally county owned bridges are in relatively worse condition
than state-owned bridges. To understand local counties’ bridge management process and identify
current practices, 50 select county maintenance personnel in the state of Georgia are interviewed.
Additionally, bridge element inspection data, reports, and deficiency characteristics are analyzed.
The goal of this study is to identify a strategic prioritization process in asset management ofcounty owned bridges that are inspected by the state department of transportation. It is concluded
that more resources have been allocated to address substructure elements and erosion issues as
they are perceived to be more vulnerable, whereas small preventive maintenance items such as
joint sealing and steel painting have been delayed. The findings indicate that county-owned steel
bridges and elements are vulnerable because local governments do not generally engage inbridge painting. It is also concluded that deficiencies are consistently indicated in inspection
reports over the years; however, inaction persists until counties are informed that bridges need to
be posted or even worse, closed. The effect of inaction is presented with real-life examples, andthe county survey and data analysis results are described with respect to a strategic prioritization
process required for managing locally-owned bridges.
1. Introduction
The Georgia Department of Transportation inspects locally-owned bridges in 159 counties andsends a deficiency report to each county. However, the National Bridge Inventory Element(NBIE) data indicate that inaction is prevalent, forcing a replacement or major repair of bridges
in less than 50 year of their 75-year service life.
2. Current Bridge Asset Management PracticeCurrently, the Georgia Department of Transportation inspects locally-owned bridges andprovides a list of maintenance items indicating deficiencies with three-tier priority ratings of A,
B, and C, representing ‘Immediate maintenance required-hazard to public safety’, ‘Schedulemaintenance’, and ‘Preventative maintenance’, respectively. While the element inspection data
provide a flagging mechanism for identifying critical bridge elements in additional to the original
National Bridge Inventory components (deck, superstructure, and substructure), descriptive
information identifying vulnerable bridge elements is obtained in inspection reports whichcurrently are available as pdf files, including photos taken during an inspection. Unfortunately,
most local governments do not always fully review the inspection reports but mainly rely on
deficiency reports, focusing on actionable maintenance items provided in a summary letter fromthe state agency.

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2.1. Analysis of the Current Practices
Bridges in the state of Georgia are in relatively good condition in the U.S. However, once
substructure concrete elements develop cracks, they rapidly develop into section loss andspalling. Thus, it is reasonable to prioritize maintenance of substructure elements, particularly
those in close proximity to ocean waterways. On the other hand, joints on the deck are notalways repaired when minor damage or small spalling is observed. However, signs of associateddeterioration are observed in substructure components such as pile caps, columns, pier caps, andabutments due to the lack of preventive maintenance conducted on joints. Additionally,
protective coating for steel beams is considered as preventive maintenance and thus has not beenprioritized, resulting in corrosion and section loss and a bridge closure when no action is taken
by county governments.
2.2. Three Most Vulnerable Bridge Components and Associated Elements
Table 1 shows four selected bridges for illustration, their three main components, and associatedbridge elements. Approximately, 100 bridges are identified by a reliability analysis conductedusing field measured vehicle weight information (Sinha et al., 2022). The change in theoperating load rating shown below is determined from the NBI database, and the health indices
are calculated by using nonlinear weight factors (Oyegbile et al., 2021). Finally, descriptive
information in the inspection reports from years 2012 to 2022 are reviewed to search for
additional details on deficiency items.
Table 1. Selected Bridge Structures, Element Health Index, and Qualitative Data.
BridgeNumberUsed forIllustration

OperatingRatingReduction(Year Built)

Reduction in Health Index > 5%

Areas of Concern in Inspection Reports

Deck &
jointsSuper
structureSub
structure

Deck
Steel
grating,
Joints,Rail

RC beams,
Steel truss

RC Deckdeterioration

Steel girder

172%(2016)-

Steel Truss

RC Column,
RC PierWall,
RC Pile2

63%(2009)-

Steel
Girder/
Beam

-

3

61%
(2009)RC TopFlange

-

-

4

58%
(2006)

Pourable
Joint-

PSC Pile

RC Deck,
Approachslabs
RC Deck,
Joints,Approachslabs,
GuardrailSuper
structure

PSC beams,
RC beams
PSC beams,
associatedbearingSubstructure
RC Pile,Column, Pier
Wall,
Pier Cap,
and Abutment
RC Column,
Abutment,
Pier CapRC Abutment,
RC Pier Cap,
PSC Pile
RC Pier Cap,
PSC Pile,
RC Abutment

3. Comparison of Quantitively and Qualitative Information
Fig. 1 through Fig. 3 present the three most frequent deficiency types found in locally owned
bridges’ inspection reports, which are public information. The quantitative information provided
in the bridge element data is compared with the more descriptive information available in bridge
inspection reports. The deterioration on the substructure was not isolated but often initiated fromdeterioration of joints on bridge deck elements. The description of deficiencies presented in thispaper includes the opinions of the author, not the state and county governments.

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(a)

(b)

(c)
(d)
Fig. 1. Deck elements of Bridges #1-4: (a) approach (b) pourable joints (c) and (d) deck joints(GDOT, 2022).

(a)

(b)

(c)
(d)
Fig. 2. Superstructure elements of Bridge #1-4: (a) steel truss section loss (b) steel truss
corrosion (c) loss of protective coating (d)steel girder ends (GDOT, 2022).

(a)

(b)

(c)
(d)
Fig. 3. Substructure elements of Bridges #1-4: (a) RC pile; (b) RC pile; (c) column; (d) pile cap
deteriorated (GDOT, 2022).

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3.1. Analysis of Interviews with Local Governments
Only 8% of the interviewed counties reported having a written asset management plan for theirbridges. Counties are heavily reliant on the state/federal support for bridge assets. Most indicate
inspection reports are challenging to understand and thus rely mainly on deficiency reports andactionable maintenance items. Counties report difficulties in finding funding for themselves
when it pertains to major repairs, compounded by ongoing difficulty locating competentcontractors. More than 50% state that they document a record of maintenance activities on theirbridges. Most common activities include brush/tree cutting, erosion control, guardrail repair, anddeck repair. Structural repair of members is cited as one of the least common.
3.2. Analysis of the NBI Condition Ratings, Element States, and Inspection Reports
Most counties do not employ a civil engineer in their public works and thus find jargon whichappears in bridge inspection reports difficult to understand. As shown in Table 1, the inspection
reports provide additional qualitative information including the detailed description of
deficiencies and interactive aspects (Oyegbile et al., 2021) of elements (e.g., joints, bearings, andsubstructure). Despite well documented deterioration particulars in the inspection reports, theinformation has not been translated to effective maintenance actions.
4. Conclusions and Future WorkA strategic prioritization process should involve budgetary planning based on prioritized
maintenance actions identified through bridge inspection and thus should include a plan forpreventive maintenance which extends the service life of bridges. Due to limited resources,
however, most local governments focus on addressing most urgent maintenance items in theirinventory. It is concluded that inspection reports provide superior qualitative information of thebridge conditions needed for maintenance whereas the quantitative element data are not as
helpful in identifying preventive maintenance items that extend service life. Such relationship isin the process of being identified by means of deep learning and other decision algorithms.
Additional challenges include the absence of photos showing bridge elements in good condition
for a comparative analysis. A priority rating of ‘B+’ requiring a maintenance action within 12
months is currently being implemented. More importantly, the largest gap between maintenanceactions and prioritized recommendation exists due to the lack of budgetary planning. A costbenefit analysis of the actionable maintenance items is not always feasible at a local-government
level, which means decision making is exclusively based on availability of funding.
5. References
GDOT, the Georgia Department of Transportation, 2022:
https://www.dot.ga.gov/Pages/Default.aspx
Oyegbile, O.B., Chorzepa, M.G., Durham, S.A. and Kim, S.S., 2021. Novel Prioritization
Mechanism to Enhance Long-Term Performance Predictions for Bridge Asset Management.
Journal of Performance of Constructed Facilities, 35(1), p.04020133.
Sinha, A., Chorzepa, M.G., Yang, J.J., Kim, S.S. and Durham, S., 2022. Enhancing Reliability
Analysis with Multisource Data: Mitigating Adverse Selection Problems in Bridge Monitoringand Management. Applied Sciences, 12(20), p.10359.

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Platform and Bigdata Driven Bridge Smart Maintenance Techniques
Ki-Tae Park1*, Kyusan Jeong2, Jaehwan Kim3, Kun-Soo Kim4, Byeongcheol Kim5, Dongwoo Seo6
1

: Korea Institute of Construction and Building Technology (KICT), South Korea, ktpark@kict.re.kr: KICT, South Korea, jungkyusan@kict.re.kr
3: KICT, South Korea, jaehwankim@kict.re.kr
4: KICT, South Korea, kunsookim@kict.re.kr
5: KICT, South Korea, bckim@kict.re.kr
6: KICT, South Korea, dwseo@kict.re.kr
2*: corresponding author

Keywords: platform, bridge, maintenance, smart
Abstract: As of December 2021, about 37,218 bridges in Korea are currently in use, of which
about 5,691 bridges are over 30 years old (15.3% of all bridges), and the number of bridges
between 20 and 30 years old is 11,820 (31.8% of all bridges). By 2032, about 47% of all bridges
are expected to be older than 30 years (MOLIT, 2022). Therefore, it is necessary to takepreventive maintenance actions by predicting the level of damage or deterioration that may occur
in the future using various data related to the current state of the bridge. Through this, the effectof reducing the huge maintenance budget that may be required in the future can be expected. In
this paper, we presented a preventive maintenance technique for bridges based on the techniqueof accumulating more than 5 million various data related to bridge deterioration and predicting
the future of bridges using the built data. Aging data includes on-site data on 100 bridges using
the IoT sensor system, long-term surveyed salinity data considering environmental conditions in
Korea, freezing and thawing, and frequency of use of deicing agents. In addition to the dataobtained in this way, unknown data was estimated using artificial intelligence techniques fornecessary but impossible data. Using the aging curve against time history, the reliability of theelement technology for evaluating the aging of the bridge based on future prediction was secured.
The accuracy of the future prediction-based algorithm was about 90.8%, which is expected toincrease as the number of data increases. The platform under development is expected to be open
after 2023, allowing users involved in bridge management to utilize it.
1. Introduction
This research aims to build more than 5 million bridge aging related data (deterioration data,
environmental data, bridge behavior, long-term survey, etc.) to advance aging bridge structure
maintenance technology. In addition, it includes the development of a smart maintenanceplatform that provides various services required for maintenance of bridge structures by user
type while securing the reliability of bridge aging evaluation element technology using artificial
intelligence techniques (bridge maintenance priority, budget basis for maintenance requirements,
aging level etc.). And also, it is to support inspection and diagnosis-related tasks currently inprogress in Korea (MOLIT, 2021).

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This research started in 2021 and is scheduled to end in 2023, and data integration and algorithm
updates will continue. In this paper, IoT sensor system installation, various data generation partusing artificial intelligence, and platform concept design results is briefly outlined. The research
results for each specific item will be presented as an additional thesis.
2. Data Integration
In this study, various data have integrated for the evaluation of aging based on the future
prediction of bridges. The integrated data results are used to provide information for preventive
maintenance through the data analysis process. This section summarizes bridge behavior dataacquisition process using IoT sensors and data integration status of Korea's environmental
condition.
2.1 IoT sensor system installation
To secure long-term data on the actual bridge behavior history, bridge aging data have integratedusing an IoT sensor system. As sensors, an accelerometer, a crack gauge, an expansion joint
gauge, an inclinometer, and a thermometer that can use IoT communication were applied. For
three years from 2021 to 2023, it was decided to install related system for a total of 100 bridges
and secure related data.
Among 954 representative bridges in Korea based on deterioration environment, it was selected
first in consideration of the year of completion and the operating environment through fieldsurveys. For sensor production and sensor installation, sensor requirements for each collection
item were reviewed (basic device configuration, operating conditions, communication conditions,etc.), and IoT sensor site installation and power/communication-related equipment installationfor each bridge were completed. The conceptual diagram related to data transmission andreception is shown in the figure below.

Fig. 1. IoT Sensor System Installation Concepts for Bridge

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2.2 Investigation of environmental conditions in Korea
In Korea, environmental conditions vary widely according to regional conditions (JS Lee et al,2010). Therefore, the environmental conditions were classified into a total of 108 cases bydividing into deicing agents use, freezing and thawing, and airborne salinity.
As of December 2022, 101 cumulative
field survey measurement points in Korea
have been selected to secure actual
measurement data of the deteriorating
environment. Among them, 60 points
were selected to investigate theenvironmental conditions for airborne
salinity along the coast, 26 points for theenvironmental for deicing agents, and 15
points for environmental surveys forFig. 2. Environmental conditions in Korea
each member of the bridge. A total of 1,300 data were obtained by the end of December 2022 inrelation to the survey of coastal airborne salinity deterioration environments.
3. Aging Level of Bridge Evaluation by Generated Data Using Artificial Intelligence
In relation to bridge structure aging data, which is difficult to collect, a study was conducted tocreate a time-series degradation curve based on collected bridge aging information and artificial
intelligence techniques. The previously obtained data was first classified by deterioration index
in consideration of the service life of the bridge, and preprocessing work for artificial
intelligence learning (classification by region and bridge type, correlation analysis, datareliability evaluation, standardization considering data characteristics etc.) was performed. And
artificial intelligence learning was used to generate time series degradation curves byregion/absence/deterioration index. In regions where data were not secured, data learning was
performed using various degradation data (weather data and degradation environment data) of
the target region to generate time-series degradation curves for each degradation index forregions where degradation data were not secured. In this research, an aging curve was createdbased on degradation curves by region, bridge type and member through artificial intelligencebased degradation curves. As a result of applying the aging prediction algorithm to six randombridges in Seoul, Korea, the accuracy of the algorithm was calculated to be 90.8% with an error
rate of 9.2% for the aging of the bridge.

(a)(b)
Fig. 3. Deterioration level evaluation accuracy: (a) bridge deck-91.5%; (b) bridge total accuracy90.8%

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4. Summary of Platform Function
The bridge smart maintenance platform developed inthis study has the function of providing information
related to bridge deterioration level evaluation andpreventive maintenance methods based on future
prediction. In other words, by using a platform on
which bridge aging data and algorithms are built, it ispossible to receive information related to the future
prediction-based aging level and preventive
maintenance plan for the bridge through the data
analysis process. In addition, the main purpose is to Fig. 4. Cover page of platform systemprovide various information related to bridge
maintenance, and a methodology that can be used as a reference along with the evaluation ofload capacity of bridges by conventional vibration measurement is also included (Soojin Cho etal, 2007).
5. Conclusion
This study was conducted to secure the reliability of bridge aging evaluation core technology byusing artificial intelligence to advance aging bridge structure maintenance methodologies.
It is necessary to take preventive maintenance actions by predicting the level of damage ordeterioration that may occur in the future using various data related to the current state of thebridge. Through this, the effect of reducing the huge maintenance budget that may be required in
the future can be expected. In this paper, we presented a preventive maintenance technique forbridges based on the technique of accumulating more than 5 million various data related tobridge deterioration and predicting the future of bridges using the built data. Aging data includes
on-site data on 100 bridges using the IoT sensor system, long-term surveyed salinity dataconsidering environmental conditions in Korea, freezing and thawing, and frequency of use of
deicing agents. In addition to the data obtained in this way, unknown data was estimated using
artificial intelligence techniques for necessary but impossible data. Using the aging curve against
time history, the reliability of the element technology for evaluating the aging of the bridge basedon future prediction was secured. The accuracy of the future prediction-based algorithm wasabout 90.8%, which is expected to increase as the number of data increases. The platform underdevelopment is expected to be open after 2023, allowing users involved in bridge management toutilize it.
6. References
Ministry of Land, Infrastructure and Transport. 2021. Detailed guidelines for safety andmaintenance of facilities (Safety Inspection and Diagnosis). KALIS
Ministry of Land, Infrastructure and Transport. 2022. Road Bridges and Tunnels Status Report.
Road bridge and tunnel status information system.
Jongseok Lee et al. 2010. Characteristics of airborne salinity distribution in Korea. Korean
Concrete Institute, 22(6), 769-776.
Sujin Cho et al. 2007. Evaluation of load carrying capacity of bridges by vibration measurement.
Proceedings of the Korean Society of Civil Engineers, 27(1), 79-89.

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