Geotechnique, Construction Materials& Environment

EDITORS
Zakaria Hassain
Suksun Horpibulsuk

ISBN: 978-4-9905958-6-9 C3051

November 14 – 16, 2016
SwissOtel Le Concorde, Bangkok, Thailand

GEOMATE 2016 BANGKOK, THAILAND
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENTiPROCEEDINGS OF SIXTH INTERNATIONAL CONFERENCE – GEOMATE 2016
GEOTECHNIQUE, CONSTRUCTION MATERIALS AND ENVIRONMENT, BANGKOK,THAILAND 14-16 NOVEMBER, 2016

Geotechnique, Construction
Materials and Environment
Edited byProf. Zakaria Hossain
Department of Environmental Science and Technology
Graduate School of Bioresources
Mie University, Japan

Prof. Suksun HorpibulsukCenter of Excellence in Innovation for Sustainable Infrastructure Development
School of Civil Engineering
Suranaree University of Technology, Thailand

THE GEOMATE INTERNATIONAL SOCIETY

ii

GEOMATE- Bangkok, Nov. 14-16, 2016

Copyright @ 2016 by The GEOMATE International Society
All rights reserved. In principle, no part of this publication or the information contained herein may be
reproduced in any form or by any means, translated in any language, stored in any data base or retrieval
system, or transmitted in any form or by any means without prior permission in writing from the publisher.
Disclaimer: The editors and the publisher have tried their best effort to ensure the integrity and the quality ofthis publication and information herein. However, they give no warranty of any kind, expressed or impliedwith regard to the material contained in this book, and will not be liable in any event for the consequences of
its use.
Published by:
The GEOMATE International Society
Tsu city, Mie, Japan
E-mail: society@geomate.org
http://www.geomate.org/

ISBN Number: 978-4-9905958-6-9 C3051

iii

Sixth International Conference on Geotechnique, Construction Materials andEnvironment Bangkok, Thailand, 14-16 Nov. 2016 ISBN: 978-4-9905958-4-5 C305

Table of Contents

PrefaceOrganization

xii
xiii

ID Keynote Papers

1

1k

SUSTAINABLE STABILIZATION OF MARGINAL LATERITIC SOIL FOR PAVEMENT APPLICAITONS
Suksun Horpibulsuk, Jeerapan Donrak, Itthikorn Phummiphan, Arul Arulrajah and Menglim Hoy

22k

ESTIMATION OF DYNAMIC BUCKLING STRENGTH OF CIRCULAR TUBE PILES WITH LIQUEFIED SOIL
Moeko Matoba, Tenshiro Goto and Yoshihiro Kimura

10

Technical Papers

20

ID

Geotechnique

21

6501

STABILITY CHARTS FOR UNSUPPORTED PLAIN STRAIN TUNNEL HEADINGS IN UNDRAINED CLAY
Jim Shiau, Brian Lamb, Mathew Sams, Jay Lobwein and Tin Nguyen

22

6508

STRENGTH AND PERMEABILITY CHARACTERISTICS OF ROAD BASE MATERIALS BLENDED WITH FLY
ASH AND BOTTOM ASH
Jonathan R. Dungca and Julie Ann L. Jao

28

6513

DISCRETE ELEMENT MODELLING OF GEOCELL-REINFORCED SUB-BALLAST SUBJECTED TO CYCLICLOADING
Ngoc Trung Ngo, Cholachat Rujikiatkamjorn and Buddhima Indraratna

34

6522

ANALYSIS OF DCM WALL SUPPORTED EXCAVATION IN CLAY BY GROUND SPRING MODEL
Siriwan Waichita, Pornkasem Jongpradist and Chanchai Submaneewong

40

6523

PROBABILISTIC SAFETY ASSESSMENT OF ANCHORED SLOPE
Takashi Hara, Akira Tomioka, Nobuyuki Soga and Atsushi Yashima

46

6524

NEW CONCEPT FOR REASONABLE GEO-DISASTER RECOVERY
Takashi Hara, Akiho Inoue, Atsushi Yashima, Nobuyuki Soga and Yoshihiro Yokota

52

6525

THE SOIL BLOCK APPROACH FOR VERIFYING HYD USING FEM
Georgios Katsigiannis, Pedro Ferreira and Raul Fuentes

58

6527

IMPROVEMENT OF HIGH EXPANSIVE SOILS BY DEEP SOIL MIXING METHOD IN THE SMALL SCALELABORATORY EXPERIMENT
Yulvi Zaika and Arief Rachmansyah

64iv

GEOMATE- Bangkok, Nov. 14-16, 2016

6529

3D-FE ANALYSES OF PILED RAFT FOUNDATION WITH CONSOLIDATION IN BANGKOK SUBSOIL
CONDITION
K. Watcharasawe, P. Kitiyodom and P. Jongpradist

71

6531

A STUDY ON PERMEABILITY OF SILICA MICRO-PARTICLES
Kentaro Uemura and Yukoh Hasunuma

77

6533

DOUBLY ASYMPTOTIC OPEN BOUNDARY CONDITION FOR MODAL RESPONSES OF PORE WATERPRESSURE
Suriyon Prempramote

83

6535

3D FINITE ELEMENT ANALYSIS OF EARTH PRESSURE BALANCE SHIELD TUNNEL EXCAVATION USINGSHELL ELEMENT AND GROUTING LAYER
P. Lueprasert, P. Jongpradist, N. Heama, K. Ruengwirojjanakul and S. Suwansawatt

92

6536

ANALYSIS OF TUNNELING CONSTRUCTION WITH GROUND-SPRING MODEL
P. Chaipanna , P. Jongpradist and M. Sugimoto

98

6538

THERMO-MECHANICAL ANALYSIS OF GEOTHERMAL PILES IN DENSE SAND
Anant Aishwarya Dubey, Suresh Kumar

103

6539

MODEL TEST ON THE REINFORCING EFECT OF THE FACE BOLTS IN THE UNCONSOLIDATED SANDYLAYER TUNNEL
Yoshifumi Taguchi

111

6542

TUNNEL LINING RESPONSES DUE TO ADJACENT LOADED PILE – NUMERICAL INVESTIGATION
N. Heama, P. Jongpradist, P. Lueprasert and S. Suwansawatt

117

6543

LIFE CYCLE ASSESSMENT ON LEAKAGE RISK FROM CONTAINMENT STRUCTURES AT COASTALLANDFILL SITES
Misato Sekitani, Shinya Inazumi, Suttisak Soralump and Ken-ichi Shishido

123

6544

6545

6546

ANALYTICAL EVALUATION OF THE INFLUENCE OF HOLES FORMED BY PULLING OUT PILE
FOUNDATIONS ON THE MECHANICAL CHARACTERISTICS OF THE SURROUNDING GROUND
Kazuki Nosho, Shinya Inazumi, Suttisak Soralump, Shuuichi Kuwahara and Yusuke Torigoe
DEVELOPMENT OF SILICA-BASED SOLIDIFICATION MATERIALS MADE FROM INORGANIC SOLIDWASTES
Yuuya Nakase, Shinya Inazumi, Suttisak Soralump, Takashi Shinsaka, Jun-ichi Yamazaki, Ryo
Hashimoto, Tomoyuki Mizuta and Yoshihiro Nakagishi
HYDRAULIC MODEL EXPERIMENTS ON DIMENSIONS OF ARMOR UNITS FOR THE RESILIENT
BREAKWATAERS
Shota Inoue, Kiyonobu Kasama, Mitsunari Hirasawa

129
135
139

6547

EVALUATION OF POROSITY IN BIOGROUTED SAND USING MICROFOCUS X-RAY CT
Shumpei Mitsuyama , Kazunori Nakashima, Satoru Kawasaki

145

6550

UNIAXIAL COMPRESSION TEST OF UNSATURATED MASADO UNDER CONSTANT DEGREE OF
SATURATION CONDITION AND ITS MODELING
Yuhei Kurimoto, Seiji Kobayashi, Takashi Tsunemoto and Feng Zhang

151

6554

AN ANALYTICAL EVALUATION OF RADIAL CONSOLIDATION WITH RESPECT TO DRAIN DEGRADATION
Thanh Trung Nguyen, Buddhima Indraratna and Cholachat Rujikiatkamjorn

156

6555

A STUDY ON INJECTING MICRO BUBBLE WATER MIXED WITH CRUSHED SILICA
Yukoh Hasunuma, Kentaro Uemura
Takamitsu Sasaki, Koichi Nagao, Naoaki Suemasa and Shunsuke Shimada

162

6556

SOIL BEARING CAPACITY REFERENCE FOR METRO MANILA, PHILIPPINES
Jonathan R. Dungca, Ismael Concepcion Jr., Moises Christian Mickail Limyuen,
Terence Oliver See and Marion Ryan Vicencio

168v

GEOMATE- Bangkok, Nov. 14-16, 2016

6557

VALIDATION OF METHOD FOR ESTIMATING LONG-TIME STRENGTH OF CHEMICAL GROUTCOMPOSITION WITH HEAT CURING
Shunsuke Takiura, Naoaki Suemasa and Takamitsu Sasaki

174

6561

VERIFICATION OF ELASTO-PLASTIC MODEL FOR CEMENTTREATED SOIL BEHAVIOR UNDER TRIAXIALTENSION CONDITION
Tsutomu Namikawa

180

6562

ASSESSMENT OF PHYSICAL AND MECHANICAL PROPERTIES OF COMPOSITE TILE MADE OF WOOD
CHIPS AND CERAMIC POWDER REINFORCED WITH BAMBOO FIBERS
Hermie M. del Pilar, Marisol E. Manarin, Clevin John V. Santos and Krizza Diane M. Santos

186

6564

UNCONFINED COMPRESSIVE STRENGTH OF COMPACTED DIS-TURBED CEMENT-STABILIZED SOFT
CLAY
Mohamed Ayeldeen, Yuki Hara and Masaki Kitazume

192

6567

A MODIFIED HARDENING SOIL MODEL FOR ROCKFILL MATERIALS
R. Sukkarak, and P. Jongpradist

198

6568

6577

LONG TERM MECHANICAL AND LOAD TRANSMISSION PROPERTIES OF SOIL-CONCRETE BLOCK
Kayo Doumoto, Kiyonobu Kasama, Mitsunari Hirasawa, Kouki Zen, Zentaro Furukawa, Yuichi Yahiro,
Masato Nakamichi, Makoto Yamaguchi, Takashi Umeyama, Masaaki Katagiri, Osamu Kawahara andToshiyuki Nagano
EFFECTS OF VERTICAL WALL BARRIER DUE TO RIGID PAVEMENT DEFLECTION OF FULL SCALE 1-PILE
ROW NAILED-SLAB SYSTEM ON SOFT SUB GRADE BY COMPRESSION LOADINGS
Anas Puri, Hary C. Hardiyatmo, Bambang Suhendro and Ahmad Rifa’i

205
210

6578

AN EVALUATION OF OSMOTIC TECHNIQUE UNDER ULTRAVIOLET GERMICIDAL IRRADIATIONEXPOSURE
Mohd Yuhyi Mohd Tadza, Nurhidayah Mahazam and Snehasis Tripathy

215

6585

STABILITY ANALYSIS OF SLOPES ON CLAY SOIL FOUNDATIONS BY LIMIT EQUILIBRIUM AND FINITEELEMENT ANALYSIS METHODS
Abdul Karim M. Zein and Waleed Abdul Karim

221

6609

EFFECT OF PERIODICAL RAINFALL ON SHALLOW SLOPE FAILURES: FINITE ELEMENT ANALYSIS
Avirut Chinkulkijniwat, Somjai Yubonchit, Rajeshwar Goodary and Duc Bui Van

227

6610

EXPRESSION OF AIR PRESSURE DISTRIBUTION USING SOIL/WATER/AIR COUPLED ANALYSIS
Katsuyuki KAWAI, Takuya KAWAKATSU and Atsushi IIZUKA

233

6613

CYCLIC LATERAL RESPONSE OF MODEL PILE GROUPS FOR WIND TURBINES IN CLAY SOIL
Werasak Raongjant , Meng Jing

239

6616

6622

THE USE OF POLYURETHANE TO MAINTAIN STRENGTH OF ROAD SUBGRADE FOR FLOOD DAMAGE
CONTROL
Safawati Mohd Radzi, Abdul Naser Abdul Ghani, Muhd Shahril Nizam Ismail, Ahmad Hilmy Abdul
Hamid and Khalid Ahmad
EVALUATION OF CONSTRAIN MODULUS-BULK STRESS RELATIONS OF PAVEMENT STRUCTURE
MATERIALS BY USING CBR MOULD
Sawanya Dararat and Warat Kongkitkul

245
250

6623

RATE-DEPENDENT DEFORMATION CHARACTERISTICS OF EPS BEAD-MIXED SAND
Agawit Thaothip and Warat Kongkitkul

257

6633

GEOTECHNICAL REQUIREMENTS FOR CAPTURING CO2 THROUGH HIGHWAYS LAND
M. Ehsan Jorat1, Ben W. Kolosz, Mark A. Goddard, Saran P. Sohi, Nurten Akgun, Dilum Dissanayake
and David A.C. Manning

265

6636

COMPOSITIONAL CHARACTERISTICS OF SAND MATRIX SOILS
Choy Soon Tan, Aminaton Marto and Ali Sobhanmanesh

270

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GEOMATE- Bangkok, Nov. 14-16, 2016
6639

EFFECT OF FLY ASH ON THE STRENGTH DEVELOPMENT OF BANGKOK SOIL-CEMENT
Keeratikan Piriyakul

276

6641

DRAINED SHEAR STRENGTH OF COMPACTED KHON KAEN LOESS
R. Nuntasarn and W. Wannakul

281

6645

A CASE STUDY ON SOIL IMPROVEMENT FOR REFORMING THE BED OF A CYLINDRICAL STORAGETANK USING MICROPILES.
Mohammad reza Atrchian and Ali NisariTabrizi

286

6649

BEHAVIOR OF STEEL NAILS-REINFORCED STONE COLUMN MADE OF MIXTURE OF TYRE CHIPS AND
STONE AGGREGATESTanwee Mazumder, Nishant Neeraj and Ramanathan Ayothiraman

293

6654

INCREASE OF EFFICIENCY OF CONSOLIDATION OF WEAK WATER-SATURATED SOIL VERTICAL SANDYDRAINS AND DINAMIC LOADINGS
Usmanov R. A.

299

6658

INFLUENCE OF CLAYSTONE DETERIORATION ON SHEAR STRENGTH OF BACKFILL
Pisut Rodvinij and Pitiwat Wattanachai

303

6664

FINITE DIFFERENCE ANALYSIS OF A CASE STUDY OF VACUUM PRELOADING IN SOUTHERN VIETNAM
Cong-Oanh Nguyen, Thanh Thi Tran and Van-Tram Thi Dao

308

6669

EARTHQUAKE ATTENUATION MODELS AND ITS RESPONSES TO EARTH ZONE DAM IN UPPERNORTHERN PART OF THAILAND
Tawatchai Tanchaisawat, Sirikanya Laosuwan and Nutapong Hirano

314

6674

VIBRATION CHARACTERISTICS OF DEFORMED STONE WALLS
OF JAPANESE TRADITIONAL CASTLE
Minoru Yamanaka, Hayato Ishigaki, Katsuhiko Koizumi, Shuichi Hasegawa and Hiroyuki Araki

320

6676

DETERMINATION OF THE POST-CYCLIC YIELD STRENGTH AND INITIAL STIFFNESS OF TWO PEAT
SOILS
Adnan Zainorabidin and Habib Musa Mohamad

326

6678

PROMOTION OF ICT UTILIZATION BY ELECTRIC RESISTIVITY MANAGEMENT IN FLUIDIZATIONTREATMENT PROCESS FOR GROUND IMPROVEMENT
Yasuhide Mochida , Indra Hardi

331

6691

GROUT PROPERTIES OF MICROPILE UNDER DIFFERENT SOIL CONDITIONS
Putli Yasmeen Nadeah M. Y., Hassanel Zachary A. Julinda Hena J. and Siti Aishah M. S.

337

6694

EXPERIMENTALLY MEASUREMENT AND ANALYSIS OF STRESS UNDER FOUNDATION SLAB
M. Mohyla and K. Vojtasik and M. Stolarik and M. Pinka1 and H. Lahuta

343

6698

RELATING TRADITIONAL DYNAMIC SHEAR MODULUS TO NANOINDENTATION MODULUS OF
ASPHALT BINDERS
Rafiqul A. Tarefder, Hasan M. Faisal, Umme A. Mannan

349

6702

SLOPE FAILURE MODEL TESTS CONTROLLED AGAINST THE FAILURES BY USING A FILTERING
MATERIAL AND THEIR SIMULATIONS
Anusron CHUEASAMAT, Toshikazu HORI, Hirotaka SAITO and Yuji KOHGO

355

6717

SETTLEMENT PREDICTION OF LARGE-DIAMETER BORED PILE IN BANGKOK SOILS
Thayanan Boonyarak, Zaw Zaw Aye, Nutthapon Thasnanipan, Sahapap Supawo, Sereyroath Chea

361

6719

INFLUENCE OF NONWOVEN GEOTEXTILE ON THE HYDRAULIC RESPONSE OF MECHANICALSTABILIZED EARTH WALL
Duc Bui Van, Avirut Chinkulkijniwat, Suksun Horpibulsuk, Arul Arulrajah, Somjai Yubonchit, Artit
Udomchai, Irin Limrat, Tulasi Ram Bhattarai

367

vii

GEOMATE- Bangkok, Nov. 14-16, 2016

6720

STABILIZING A STEEP WEATHERED SANDSTONE SLOPE ALONG THE RAILROAD IN SOUTHERNGERMANY
Thomas Hangartner and Christophe Balg

373

6722

REFLECTION OF TECHNOLOGIES OF NAVIGATION IN SPACE-TIME IN THE STRUCTURE OF
ARCHEOLOGICAL OBJECTS
Alina Paranina

377

ID

Construction Materials

383

6507

ACCUMULATIVE DAMAGE OF FIBER SHEET REINFORCEMENT CAUSED BY NEGATIVE THERMAL
EXPANSION COEFFICIENT UNDER CYCLIC TEMPERATURE
Hidenori TANAKA

384

6511

UTILIZATION OF AGGREGATE QUARRY WASTE IN CONSTRUCTION INDUSTRY
Mary Ann Q. Adajar, Euclid de Guzman, Ryan Ho, Cesar Palma Jr. III, and Dennis Sindico

390

6514

SPENT COFFEE GROUND-FLY ASH GEOPOLYMER: STRENGTH ANALYSIS AS A RECYCLED ROADSUBGRADE MATERIAL
TA. Kua, A. Arulrajah, S. Horpibulsuk and YJ. Du

396

6528

OPTIMIZATION OF COMPRESSIVE STRENGTH OF CONCRETE WITH PIG-HAIR FIBERS AS FIBER
REINFORCEMENT AND GREEN MUSSEL SHELLS AS PARTIAL CEMENT SUBSTITUTE
Jayvee L. Gagan and Bernardo A. Lejano

402

6530

PULL-OUT STRENGTH OF AN EXPANSION STUD ANCHOR IN CARBON FIBER REINFORCED
CONCRETE
Gilford B. Estores and Bernardo A. Lejano

408

6534

EVALUATION ON THE EFFECT OF BAMBOO STRIPS REINFORCEMENT TO THE MECHANICAL AND
PHYSICAL PROPERTIES OF RICE HULL PARTICLE BOARD
Hermie M. del Pilar, Jecelle Elyssa E. Calip , Aldwin Christopher A. Loyola1 and Neil Patrick C. Nase

415

6551

STRENGTHENING EFFECT OF A CFRP ROD HAVING RIBS FOR A CANTILEVERED SLAB
Yusuke Kuroda, Hiroaki Hasegawa, Nobuhiro Hisabe and Isamu Yoshitake

420

6558

CONSTITUTIVE MODELING OF COAL ASH USING MODIFIED CAM CLAY MODEL
Erica Elice Saloma Uy and Jonathan Rivera Dungca

426

6565

ASSESSMENT OF CRITICAL-STATE SHEAR STRENGTH PROPERTIES OF COPPER TAILINGS
Erica Elice Saloma Uy and Mary Ann Q. Adajar

432

6601

GEOTECHNICAL PROPERTIES OF LADLE FURNACE SLAG IN ROADWORK APPLICATIONS
Farshid Maghool, Arul Arulrajah, Suksun Horpibulsuk and Yan-Jun Du

438

6620

TEMPERATURE EFFECTS ON ELASTIC STIFFNESS OF HDPE GEOGRID AND ITS MODELLING
Thitapan Chantachot, Warat Kongkitkul and Fumio Tatsuoka

443

6648

EVALUATION OF DIFFERENT FIBER REINFORCED MORTAR AS RETROFITTING MATERIALS FOR RC
COLUMNS
Bernardo A. Lejano

449

6671

DEVELOPMENT OF A NEW SOIL-CEMENT BLOCK USING PRODUCED WATER FROM OIL FIELDS: A
PRELIMNARY INVESTIGATION
Khalifa S. Al-Jabri, A. W. Hago, Mahad Baawain and Gyanendra Sthapit

455

6677

WASTE PLASTICS AS REINFORCING IN STRUCTURAL COMPOSITE MATERIALS: PROPERTIES AND
CHARACTERISTICS
Youssef Halimi, Hayat Bouchoum, Souad Zyade and Mohamed Tahiri

461

viii

GEOMATE- Bangkok, Nov. 14-16, 2016

6683

MODELLING OF CARBONATION OF REINFORCED CONCRETE STRUCTURES IN INTRAMUROS,
MANILA USING ARTIFICIAL NEURAL NETWORKRichard M. De Jesus, Joshua A. M. Collado, Jemison L. Go, Mike A. Rosanto and John L. Tan

468

6689

INFLUENCE OF 3D NUMERICAL MODEL PARAMETERS IN ANALYSES OF THE SUBSOIL-STRUCTURE
INTERACTION
Jana Labudkova and Radim Cajka

473

6712

INVESTIGATION OF POTENTIAL ALKALI-SILICA REACTIVITY OF AGGREGATE SOURCES IN THAILAND
Suvimol Sujjavanich, Krit Won-In, Thanawat Meesak, Watcharagon Wongkamjan and Viggo Jensen

479

6721

CBR BEHAVIOR OF SOFT MARINE CLAY TREATED WITH CLASS ‘F’ FLY ASH AND NANO MATERIAL
Prasanna P.Kulkarni and J. N. Mandal

485

ID

Environment

493

6532

USING FLOATING WETLAND TREATMENT SYSTEMS TO REDUCE STORMWATER POLLUTION FROMURBAN DEVELOPMENTS
Peter Schwammberger, Chris Walker and Terry Lucke

494

6548

PROPOSITION OF A DETERIORATION PREDICTION MODEL FOR MAINTENANCE OF
RUNWAY PAVEMENT
Hiroto Akimoto, Naoaki Suemasa, Kazuya Itoho, Tsuyoshi Tanaka

500

6549

ARTIFICIAL NEURAL NETWORK PERMEABILITY MODELING OF SOIL BLENDED WITH FLY ASH
Joenel G. Galupino and Jonathan R. Dungca

506

6553

ITERATIVE ALGORITHM TO CONSTRUCT AN EXACT FINITE ELEMENT MODEL FOR AXIALLY LOADEDPILE IN ELASTO-PLASTIC SOIL
Chinapat Buachart and Chayanon Hansapinyo

512

6559

ARTIFICIAL SLUDGE BASED ON COMPOSITIONAL INFORMATION OF A NATURAL SEA SLUDGE
Hirosuke Hirano, Takeshi Toyama, Nobuyuki Nishimiya, Davin H. E. Setiamarga, Shugo Morita, Yuto
Uragaki, and Kyoichi Okamoto

517

6572

COLONIZATION AND MORPHOLOGICAL CHANGES OF A SEDGE RESTRICTING REGENERATION AFTER
WIND DAMAGE IN A NATURAL FOREST IN KISO DISTRICT, CENTRAL JAPANTeruo Arase, Tetsuo Okano and Tetsuoh Shirota

522

6573

METHODS OF SUPPRESSING COLONIZING SEDGE TO HELP TO ESTABLISH TREE SEEDLINGS IN A
NATURAL FOREST IN KISO DISTRICT, CENTRAL JAPANTeruo Arase, Tetsuo Okano and Tetsuoh Shirota

526

6576

6584

6589

DYNAMICS OF EXOTIC GRASS COVER PLANTS ON SLOPES
Taizo Uchida, Jun Tanaka, Kentaro Kondo, Daisuke Hayasaka, Yuki Tomoguchi, Teruo Arase andTetsuo Okano
NEW DATA ON AFRICAN-EURASIAN LIMIT AND SPATIAL DISTRIBUTION OF THE CURRENT
DEFORMATION THROUGH ACTIVE STRUCTURES IN THE CENTRAL MEDITERRANEAN AND THE
SURROUNDING AREAS
Adnène KASSEBI , and Fouad ZARGOUNIREMOVAL OF ORGANIC MATTER IN WASTEWATERS OF A MILK FACTORY AND A HOSPITAL USING A
CUBIC LATTICE BASED ROTATING BIOLOGICAL CONTACTOR IN VIETNAM
Tatsuhide Hamasak, Phan Do Hung and Hiroshi Tsuno

532
538
545

6590

GREEN INNOVATION OF CALCIUM SULFATE OR GYPSUM PREPARATION FROM DUCK EGGSHELL VIAPYROLYSISNuchnapa Tangboriboon and Wanitcha Unjan

551

6591

VELOCITY STRUCTURE AND EARTHQUAKE RELOCATIONS AT CENTRAL PENINSULAR MALAYSIAREGION
Abdul Halim Abdul Latiff and Amin Esmail Khalil

557

ix

GEOMATE- Bangkok, Nov. 14-16, 2016

6598

VEGETATIVE INFLUENCE ON ROUGHNESS LEVELS FOR PAVEMENTS FOUNDED ON ALLUVIAL
EXPANSIVE SOIL DEPOSITS
Md Yeasin Ahmed, Robert Evans and Monzur Imteaz

563

6600

INFLUENCE OF WEATHERING OF BOTTOM ASH ON THE LEACHING BEHAVIOR OF CESIUM
Yasumasa Tojo, Saori Iwamoto, Mikako Ishii, Toshihiko Matsuto and Takayuki Matsuo

569

6602

THERMOELECTRIC POWER (TEP) MEASURMENT OF GEL GROWN BARIUM OXALATE CRYSTAL
Paresh Vasantlal Dalal

575

6603

IMPACT ON AIR QUALITY BY INCREASE IN AIR POLLUTANT EMISSIONS FROM THERMAL POWER
PLANTS
Akira Kondo, Hikari Shimadera and Mai Chinzaka

578

6604

OCEAN DECONTAMINATION: HIGH ABILITY REMOVAL METHOD TO RADIOACTIVE CESIUM FROM
OCEAN SLUDGE BY USING MICRO BUBBLES AND ACTIVATING MICROORGANISMS
Kyoichi OKAMOTO, Takeshi TOYAMA and Tomoe KOMORIYA

584

6605

EFFECT OF ADDITION OF BACTERIA ON THE REMOVAL OF RADIOACTIVE CESIUM FROM OCEANSLUDGE IN A CIRCULATION TYPE PURIFICATION SYSTEM
Tomoe KOMORIYA, Kyoichi OKAMOTO and Takashi TOYAMA

590

6611

A STUDY ON GROUND IMPROVEMENT TECNIQUE WITH IN-SITU MICROOGANISMS ISOLATEDFROM JAPAN
Rusutsu Ito, Toshiro Hata

596

6618

A STUDY OF RESTRAINT TECHNIQUES FOR CEMENT TREATED SOIL’S DETERIORATION BY
MICROBIAL FUNCTIONSKazuki Mihara, Toshiro Hata

600

6621

REMOVING FLUORIDE FROM A HOT SPRING USING AN ELECTROLYSIS SYSTEM Yuki Imai, Shiori
Yanagawa, Misa Konishi, and Tomonori Kawakami

604

6626

GENETIC DIVERSITY AND GENETIC STRUCTURE OF AN ENDANGERED SPECIES, ERIOCAULONNUDICUSPE, GROWING IN ARTIFICAL DISTURBING HABITATS
Michiko Masuda, Tadamasa Fukagawa and Fumitake Nishimura

610

6627

THE ANALYSIS OF FLOODING CONDITION USING BY GIS DATA IN THE THAILAND IN 2011
Masanobu Taniguchi, Hiroyuki Ii, Yuji Hara1 and Danai Thaitakoo

618

6643

UTILIZATION OF TOFU LIQUID WASTE GENERATED FROM ANAEROBIC PROCESSING IN COMPOST
PREPARATION
M. Faisal

623

6650

COMPARATIVE ANALYSIS OF NEW GROUND MATERIAL AND EMBANKMENT CONSTRUCTIONMETHODS IN CONSIDERATION OF RECYCLING BY LCA
Hideyuki Ito, Koichi Yamanaka , Hideo Noguchi , Takahiro Fujii and Kunio Minegishi

629

6651

TSUNAMI IMPACT ANALYSIS TO GEOLOGICAL LANDSCAPE IN PERAK COAST, MALAYSIA
Sukhveender Singh Sukhdev Singh and Abdul Halim Abdul Latiff

635

6653

THE EXPERIMENTAL DESIGN AND CARBON FOOTPRINT ASSESSMENT OF NON-GLAZED FLOOR TILES
Karin Kandananond

640

6682

DAMAGE QUANTIFICATION OF BEAMS USING FREQUENCY SIGNATURE
Bryan Josef T. Medrano and Lessandro Estelito O. Garciano

646

6690

STUDY DOWNTOWN STVANGER TO THE PEDESTRIAN SPACE
Tetsuya Yanobe, Kazunari Tanaka and Shin Yoshikawa

652x

GEOMATE- Bangkok, Nov. 14-16, 2016

6693

ON THE RELATIONSHIP BETWEEN THE SPATIAL ELEMENTS AND PERCEPTUAL SPACE OF CHILDREN
Takumi Sakai, Kazunari Tanaka and Shin Yoshikawa

656

6699

CHARACTERISTICS OF COLLAPSE TIME OF EARTHQUAKE SOURCE ROCK AS ONE PARAMETER TO
PREDICTION THE EARTHQUAKE IN YOGYAKARTA REGION
Hita Pandita, Sukartono and Agustinus Isjudarto

660

6704

MATHEMATICAL MODELING OF FOREST FIRES INITIATION, SPREAD AND IMPACT ON
ENVIRONMENT
Valeriy Perminov, Alexander Goudov

665

6715

UNIT WEIGHT AND COMPRESSIVE STRENGTH OF CELLULAR LIGHTWEIGHT RECYCLED GLASS-FLY
ASH GEOPOLYMER
Cherdsak Suksiripattanapong, Rutchakit Maythathirut, Sermsak Tiyusangthong, Suksun
Horpibulsuk and Arul Arulrajah

671

6686

FINITE ELEMENT SIMULATION OF EMBANKMENTS UTILIZING SMALL-SCALE CENTRIFUGALDIMENSIONSAli Sobhanmanesh, Ramli Nazir, Tan Choy Soon and Deprizon Syamsunur

675

6582

BEHAVIOR OF ACCIDENTAL CASES OF TBM SEGMENTAL LINING
Mostafa Zaki, Ahmed Hassan and Mostafa Asaad

681

6668

EFFECT OF THICKNESS ACOUSTIC PANELS UTILIZING COCONUT COIR
D.Salinah, M.T.Fadzlita, G.Habibah, L.Janice Ayog, A.Adriana and A.Hassanel

687

Authors indexxiSixth International Conference on Geotechnique, Construction Materials andEnvironment Bangkok, Thailand, 14-16 Nov. 2016 ISBN: 978-4-9905958-4-5 C305

PrefaceOn behalf of the GEOMATE 2016 Organizing Committee, we would like to welcome you in attending theInternational Conference on Geotechnique, Construction Materials and Environment held at the Swissotel Le
Concorde, Bangkok, Thailand in conjunction with the GEOMATE International Society, Southeast Asian
Geotechnical Society (SEAGS), Suranaree University of Technology, Thailand, AOI-Engineering, Useful
Plant Spread Society, HOJUN, Cosmo Winds, and Glorious International, JapanOn Friday 11 March 2011, at 14:46 Japan Standard Time, the north east of Japan was struck and severely
damaged by a series of powerful earthquakes which also caused a major tsunami. This conference was first
dedicated to the tragic victims of the Tohoku-Kanto earthquake and tsunami disasters. The Geomate 2016conference covers three major themes with 17 specific themes including:
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Advances in Composite Materials
Computational Mechanics
Foundation and Retaining Walls
Slope Stability
Soil Dynamics
Soil-Structure Interaction
Pavement Technology
Tunnels and Anchors
Site Investigation and Rehabilitation
Ecology and Land Development
Water Resources Planning
Environmental Management
Public Health and Rehabilitation
Earthquake and Tsunami Issues
Safety and Reliability
Geo-Hazard Mitigation
Case History and Practical Experience

This year we have received many paper submissions from different countries all over the world, including
Australia, Bank Receipt, Czech Republic, India, Indonesia, Japan, Malaysia, Oman, Philippines, Russia,
Switzerland, Tajikistan, Thailand, Tunisia, United Kingdom, United States and Vietnam. The technicalpapers were selected from the vast number of contributions submitted after a review of the abstracts. The
final papers in the proceedings have been peer reviewed rigorously and revised as necessary by the authors. It
relies on the solid cooperation of numerous people to organize a conference of this size. Hence, we
appreciate everyone who support as well as participate in this joint conferences.
Last but not least, we would like to express our gratitude to all the authors, session chairs, reviewers,
participants, institutions and companies for their contribution to GEOMATE 2016. We hope you enjoy theconference and find this experience inspiring and helpful in your professional field. We look forward toseeing you at our upcoming conference next year.
Best regards,
Prof. Dr. Suksun Horpibulsuk, Conference Chairman
Prof. Dr. Md. Zakaria Hossain, Conference Directorxii

Sixth International Conference on Geotechnique, Construction Materials andEnvironment Bangkok, Thailand, 14-16 Nov. 2016 ISBN: 978-4-9905958-4-5 C305

Organization
Conference Honorary Chairmen:
Emeritus Professor Dr. Sohji Inoue, Mie University, Japan
Emeritus Professor Dr. Teruo Nakai, Nagoya Institute of Technology, Japan
Emeritus Professor Dr. Fusao Oka, Kyoto University, Japan
Scientific Committees:
Conference Chairman: Dr. Suksun Horpibulsuk, Prof. Suranaree University of Tech., Thailand
Conference Vice Chairmen: Dr. Avirut Chinkulkijniwat, A/Prof. Suranaree Univ of Tech., Thailand
Dr. Jim Shiau, Senior Lecturer, USQ, Australia
Conference Director: Dr. Zakaria Hossain, Prof. Mie University, Japan
Conference General Secretary: Dr. Shinya Inazumi, A/Prof. National Ins. of Tech., Akashi, Japan
Conference Organizing Committee:
Dr. Suksun Horpibulsuk, Prof. Suranaree Uni. of Tech., (Chair)
Dr. Watcharin Gasaluck, Prof. Khon Kaen Uni., (Co-Chair)
Dr. Avirut Chinkulkijniwat, A/Prof. Suranaree Univ of Tech. (VC)
Dr. Jim Shiau, Senior Lecturer, USQ, Australia (VC)
Dr. Sohji Inoue, E/Prof. Mie University, Japan (Co-Chair)
Dr. Toshinori Sakai, Prof. Mie University, Japan (Co-Chair)Dr. Zakaria Hossain, Prof. Mie University, Japan (Director)
Dr. Shinya Inazumi, A/Prof. National Ins. of Tech., Akashi (GS)National & International Advisory Committee:
Dr. Sittisak Uparivong, Prof. Khon Kaen University, Thailand
Dr. Jirayut Suebsuk, A/Prof. Rajamangala Uni of Tech., Thailand
Dr. Phacharaporn B., A/Prof., Kasetsart Uni., Thailand
Dr. Dolrerdee Hormdee, A/Prof., Khon Kaen Uni, Thailand
Dr. Vanchai Sata, A/Prof., Khon Kaen University, Thailand
Dr. Chinapat Buachart, A/Prof. Chiang Mai University, Thailand
Dr. Mark Jaksa, Prof., University of Adelaide, Australia
Dr. Fumio Tatsuoka, Prof. Tokyo University of Science, Japan
Dr. Jing-Cai Jiang, Prof. University of Tokushima, Japan
Dr. Toshihiro Morii, Prof. Niigata University, Japan
Dr. Kimitoshi Hayano, Prof. Yokohama National Univ., Japan
Dr. Sai Vanapalli, Prof. University of Ottawa, Canada
Dr. M. Bouassida, Prof. National Sch. of Engg. of Tunis
Dr. L.R. Austriaco, Prof. Angles Univ. Found, Philippines
Dr. A.S.M. Abdul Awal, Prof. Univ. Technology Malaysia
Dr. M. Ibn Ibrahimy, Prof. Int. Islamic Univ., Malaysia
Dr. Bujang B.K. Huat, Prof. Univ. Putra Malaysia
Dr. Nemy Banthia, Prof. UBC, Canada
Dr. Ian Jefferson, Prof. Univ. of Birmingham, UK
Dr. John Bolander, Prof. Univ. of California, USA
Dr. Shamsul Chowdhury, Prof. Roosevelt Univ., USA
Dr. Isabel Pinto, Prof. University of Coimbra, Portugal
Dr. Hj. Ramli Bin Hj. Nazir, A/Prof.. UTM, Malaysia

xiiiGEOMATE- Bangkok, Nov. 14-16, 2016
Dr. Aly Ahmed, A/Prof..Beni-Suef University, Egypt
Dr. Chang-Yu Ou, Prof. National Taiwan Univ. of Sci. &Tech.
International Technical Program Committee:
Prof. Sai Vanapalli, University of Ottawa, Canada
Prof. Alaa Masoud, Tanta University, Egypt
Prof. Aly Ahmed, Beni-Suef University, Egypt
Prof. Aminaton Marto, Universiti Teknologi Malaysia, Malaysia
Prof. Bandari Shankar, Osmania University, India
Prof. Bashir Ahmed Mir, National Institute of Technology, India
Prof. Hamidi Abdul, Aziz, Universiti Sains Malaysia (USM), Malaysia
Prof. Hussein Elarabi, University of Khartoum, Sudan
Prof. M. Shohidullah Miah, International Univ. of Business Agriculture &Tech., Bangladesh
Prof. Michele, Casagrande, Pontifical University Catholic of Rio De Janeiro, Brazil
Prof. Radim Cajka, Technical University of Ostrava, Czech RepublicProf. Rajaraman Jambunathan, AMET University, India
Prof. Reshma Chandran T., Global Institute of Architecture, India
Prof. Roslan Hashim, University of Malaya, Malaysia
Prof. Seyed Naser Moghaddas Tafreshi, K.N. Toosi University of Technology, IranProf. Valeriy Perminov, Tomsk Polytechnic University, Russia
A/ Prof. Abdoullah Namdar, Sichuan University, China
A/ Prof. Abdul Naser Abdul Ghani, Universiti Sains Malaysia, Malaysia
A/ Prof. Alaeddinne Eljamassi, Islamic University of Gaza, Palestine
A/ Prof. Alina Paranina, Herzen State Pedagogical University of Russia, Russia
A/ Prof. Ashraf Elmoustafa, Ain Shams Univeristy, Egypt
A/ Prof. Bindu C S, Cochin University of Science& Technology, India
A/ Prof. Chee-Ming Chan, Universiti Tun Hussein Onn Malaysia, Malaysia
A/ Prof. Chidanand Naik, Anjuman Instistute of Technology and Management, India
A/ Prof. Dahlia Hafez, Cairo University, Egypt
A/ Prof. Deepa G Nair, Isfahan University of Technology, India
A/ Prof. Dolrerdee Hormdee, Khon Kaen University, Thailand
A/ Prof. Farhad Behnamfar, Isfahan University of Technology, Iran
A/ Prof. Gabriela B. Cazacu, Geotech Dobrogea, Romania
A/ Prof. Hadi Khabbaz, University of Technology, Sydney (UTS), Australia
A/ Prof. Homayoon Ganji, Herat Univercity Badghis, Afghanistan
A/ Prof. Hudson Jackson, US Coast Guard Academy, United States
A/ Prof. Inazumi Shinya, National Institute of Technology, Akashi, Japan
A/ Prof. John Smith, RMIT University, Australia
A/ Prof. Kasinathan Muthukkumaran, National Institute of Technology, India
A/ Prof. Lindrianasari Lindrianasari, Universitas Lampung, Indonesia
A/ Prof. Mahdi Karkush, Baghdad University, Iraq
A/ Prof. Mohamed Redha Menani, Batna University, Algeria
A/ Prof. Navid Khayat, Ahvaz Branch, Islamic Azad University, Iran
A/ Prof. Salam, Bash AlMaliky, AlMustansiriya University, Iraq
A/ Prof. Shailesh Kumar Jha, Indian Institute of Technology, India
A/Prof. Teodor Lucian Grigorie, University of Craiova, Romania
Dr. Abdul Karim M. Zein, University of Khartoum, Sudan
Dr. Abdull Halim Abdul, Universiti Teknologi MARA, Malaysia
Dr. Afshin Asadi, University Putra Malaysia, Malaysia
Dr. Ahmad Safuan A Rashid, Univeristi Teknologi Malaysia, Malaysia
Dr. Ahmad Safuan Bin A Rashid, Univeristi Teknologi Malaysia, Malaysia
Dr. Akindele Okewale, Federal University of Petroleum Resources, Nigeria
Dr. Akinola Johnson Olarewaju, Federal Polytechnic Ilaro, Nigeria
Dr. Ali Sobhanmanesh, Universiti Teknologi Malaysia (UTM), Malaysia
Dr. Alireza Bahiraie, Semnan University, Iran
Dr. Allan Manalo, Centre of Excellence in Engineered Fibre Composites, Australia
Dr. Aniza Ibrahim, National Defence University of Malaysia, Malaysia

xivGEOMATE- Bangkok, Nov. 14-16, 2016
Dr. Arif Ali Baig Moghal, King Saud University, Saudi Arabia
Dr. Aslan S. Hokmabadi, University of Technology, Sydney (UTS), Australia
Dr. Ather Ashraf, PUCIT, Old Campus Punjab University, Pakistan
Dr. Atsuko Sato, Cicil Engineering Research Institute for Cold Region, Japan
Dr. Ben-Hur Silva, Military Institute of Engineering, United States
Dr. Choy Soon Tan, Univeristi Teknologi Malaysia, Malaysia
Dr. David Thorpe, University of Southern Queensland, Australia
Dr. Delsye Ching Lee Teo, Universiti Malaysia Sarawak, Malaysia
Dr. Domenico Lombardi, University of Manchester, United Kingdom
Dr. Ehsan, Jorat Newcastle University, United Kingdom
Dr. Ganesh Kumar Shanmugam, National Institute of Ocean Technology, India
Dr. Helsin Wang, Institute of Bridge Engineering, Taiwan
Dr. Hossein Moayedi, Kermanshah University of Technology, IranDr. Hossein MolaAbasi, Babol University of Technology, Iran
Dr. James Hambleton, University of Newcastle, Australia
Dr. James Hambleton, University of Newcastle, Australia
Dr. Janaka Kumara, Tokyo University of Science, Japan
Dr. Jirayut, Suebsuk Rajamangala University of Technology Isan, Thailand
Dr. Juhyung Lee, Korea Institute of Civil Engineering and Building Technology, South Korea
Dr. Jun Sugawara, Advision - WorleyParsons Group, Australia
Dr. Lamia Touiti Bouebdellah, Ecole Nationale d'Ingénieurs de Gabes, Tunisia
Dr. Luky Handoko, Universitas Atma Jaya Yogyakarta, Indonesia
Dr. M. Mohammad Ali, California Public Utilities Commission, United States
Dr. Marfiah Ab.Wahid, Universiti Technologi Mara, Malaysia
Dr. Maryam Naeimi, Semnan University, Iran
Dr. Mehdi Mokhberi, Islamic Azad University, Shiraz, Iran
Dr. Mohd Hairy Ibrahim, Sultan Idris Education University, Malaysia
Dr. Neelima Satyam, International Institute of Information Technology, India
Dr. Nor Zurairahetty Mohd Yunus, Univeristi Teknologi Malaysia, Malaysia
Dr. Siavash Zamiran, Southern Illinois University Carbondale, United States
Dr. Subha Vishnudas, Cochin University of Science and Technology, India
Dr. Sunggi Jin, SK E&C, South Korea
Dr. Sunil Pusadkar, Govt. College of Engineering, Amravati, India
Dr. Teresa Lopez-Lara, Universidad Autonoma De Queretaro, Mexico
Dr. Usama Juniansyah Fauzi, University of Tokyo, Japan
Dr. Yusep Muslih Purwana, Sebelas Maret University, Indonesia
Conference Correspondence:Prof. Dr. Zakaria Hossain, Conference Director,
Dept. of Env. Sci. & Tech., Mie University, Japan,
Prof. Dr. Suksun Horpibulsuk (Chairman)
Dept. of Civil Engineering, Suranaree University of Technology, Thailand
Dr. Shinya Inazumi, General Secretary
A/Prof. National Ins. of Tech., Akashi, Japan
E-mail: conference@geomate.org
Tel & Fax: +81-59-231-9578
Editorial Committee and Executive Committee:
Dr. Zakaria Hossain
Dr. Suksun Horpibulsuk
Dr. Avirut Chinkulkijniwat
Dr. Jim Shiau
Dr. Sinya Inazumi

xv

GEOMATE- Bangkok, Nov. 14-16, 2016

Keynote Paper

1Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

SUSTAINABLE STABILIZATION OF MARGINAL LATERITIC SOIL
FOR PAVEMENT APPLICAITONS
Suksun Horpibulsuk1, Jeerapan Donrak2, Itthikorn Phummiphan3, Arul Arulrajah4 and Menglim Hoy5
Center of Excellence in Innovation for Sustainable Infrastructure Development, and School of Civil
Engineering, Suranaree University of Technology Thailand; 3Department of Rural Roads, Thailand; 4Department
of Civil and Construction Engineering, Swinburne University of Technology, Australia
1,2,5

ABSTRACTDue to high rainfall, temperature and humidity with alternative wet and dry period, nearly 60% of the soils in
Thailand are lateritic soil with colors ranging from red to yellowish red. The lateritic soil is found in dry flat
lands and plains, throughout Thailand. This lateritic soil with suitable mechanical properties is commonly usedas subbase materials in roads. However, lateritic soils are increasingly becoming scarce to source for roadprojects. Thus, the usage of marginal lateritic soil as a pavement subbase material leads to some challenging
issues. This paper presents sustainable stabilization of marginal lateritic soil by Melamine Debris (MD)replacement and by geopolymer binders as unbound and bound pavement materials, respectively. MD is anindustrial waste resulting from plate and cup manufacture. This MD cannot be reformed or reused for
manufacturing plates and cups. Fly Ash (FA) is used a precursor to develop geopolymer binder. The test results
on MD-lateritic soil blends shows that the MD replacement of lateritic soil reduces the fine content and increases
the abrasion resistance of the soil particles, hence the reduction in liquid limit, plasticity index, LA abrasion andparticle breakage. The physical and mechanical properties of the 20% MD replacement blend are found to meetthe requirement of local road authority for engineering fill materials while the 50% MD replacement blend isfound to be at the borderline for subbase course material. The test results on geopolymer stabilized lateritic soil
show that the maximum early strengths at 7 days of curing were found at Na 2 SiO 3 :NaOH of 90:10, where
Calcium Silicate Hydrate (C-S-H), cementitious products from high calcium FA and Na 2 SiO 3 , was found to playa significant role. The Sodium Alumino Silicate Hydrate (N-A-S-H) products, being time-dependent, howevercame into play after a longer duration. The maximum 90-day UCS was found at a Na 2 SiO 3 :NaOH ratio of 50:50.
This research is significant in addressing the sustainable usage of MD and green geopolymer binder fromengineering and environmental perspectives.
Keywords: marginal lateritic soil, geopolymer, melamine, calcium carbide residue, fly ash

resulting from plate and cup manufacture. This MD
cannot be reprocessed for manufacturing plates andcups [4]. It was reported by Srithai Superware PublicCo., Ltd, a largest melamine tableware company inThailand, that the company releases 400 tons of MD
annually. Presently, this MD is disposed by
combustion at a very high temperature, which iscostly and has significant negative environmental
effects. The sustainable usage of MD in civilengineering applications will provide a lower carbon
footprint and result in positive social and economical
impacts for governments, industry and consumers.
The usage of MD in marginal soil improvement forpavement application is innovative and of interest to
the industrial sectors and national road authorities,
particularly as road construction typically requires alarge quantity of quality material.
Another alternative way, widely practiced aroundthe world, is to compact the in-situ marginal soil
mixed with Portland Cement (PC) [5]. The
production of PC is however an energy-intensive
process and emits a very large amount of greenhouse

INTRODUCTION
Due to high rainfall, temperature and humidity
with alternative wet and dry period, nearly 60% of
the soils in Thailand are lateritic soil with colors
ranging from red to yellowish red. Quality lateritic
soil for pavement subbase/base and engineering fillmaterials are becoming increasingly scarce to source
for road infrastructure projects. Therefore,
modification of soils on site is an attractive andeconomical option to improve their mechanicalproperties [1].
The most common soil improvement technique isto compact the locally available soil blended withhigh quality materials. Compaction is a practical andeconomical technique to improve the mechanicalproperties of soils by the application of either staticor dynamic force [2]. The usage of recycledmaterials blended with high quality materials tostabilize marginal soils has significant engineering,
economical and environmental benefits [3].
Melamine debris (MD) is an industrial waste
2GEOMATE- Bangkok, Nov. 14-16, 2016

gas, particularly carbon dioxide (CO 2 ) into theatmosphere [6]. In the light of these problems, theutilization of low energy intensive cementing agentsin civil engineering applications has been
increasingly researched in recent years. Low energy
intensive cementing agents include cement kiln dust
[7] calcium carbide residue [8] granulated blast
furnace slag [9] and geopolymer binder [10].
Geopolymer is an inorganic aluminosilicate
material synthesized by alkaline activation ofmaterials rich in alumina (Al 2 O 3 ) and silica (SiO 2 ).
Since Fly Ash (FA) offers the greatest opportunity
given the plentiful worldwide stockpiles, it isfrequently used as a precursor for producing
geopolymers. Though there is readily available
literature on the application of FA based geopolymer,
they have been predominantly used as building
materials while the studies on the application of FA
based geopolymer to pavement materials are limited
to date.
The paper attempts investigate physical andmechanical properties of marginal lateritic soil
blended with MD at various replacement ratios toascertain it as an engineering fill material and toinvestigate the possibility of using high calcium FA
based geopolymer to stabilize a marginal lateritic
soil to be a sustainable bound pavement material.
This study is significant from engineering,
economical and environmental perspectives.

reduction in fine content results in the decreases in
the liquid limit (w L ) and plasticity index (I p ); i.e.,w L is 39.9%, 39.8%, 38.9%, 38.4% and 7.8% and I p
is 17.0, 16.6, 15.1, 12.6 and 11.1 for MD
replacement ratios of 10%, 20%, 30%, 40% and50%, respectively. Consequently, the lateritic
soil/MD blends approach to a silty material typewith increasing MD replacement ratio; i.e., they areclassified as SM, SM, SM, SC and SC for MD
replacement ratios of 10%, 20%, 30%, 40% and50%, respectively while the lateritic soil is classified
is SC. It is noted that w L and I p of these blends meetthe consistency limits specified for engineering fillmaterials (w L < 40% and I p < 20%).
The MD replacement technique has theadvantage of reduced particle breakage of the blends
due to compaction as evident from Fig. 1 and 2.
Besides the Atterberg limits and gradation, thedurability of the lateritic soil particles is improvedby the MD replacement. The MD has a relatively
low LA abrasion of 11.3% compared to the lateritic
soil with LA equal to 58.1%. Therefore, the MD
replacement improves the LA abrasion of thelateritic soil by reducing the LA abrasion value forthe blends. The LA abrasion of lateritic soil
decreases from 58.1% (for 0% MD replacement) to
41.1% (for 50% MD replacement). Even without the
MD replacement, this lateritic soil is considered as
durable for subbase and engineering fill materials
based on LA abrasion requirement of < 60% [11].
The modified compaction test results (Fig. 4)
show that the blends at various MD replacement
ratios exhibit bell-shaped compaction pattern,
typical of traditional geo-materials [2], [12]. The
Maximum Dry Density (MDD) values of the blends
are between those of the lateritic soil and MD, whichare 19.4 kN/m3, 18.9 kN/m3, 18.2 kN/m3 and 17.3
kN/m3 for 10%, 20%, 30% and 40% MD
replacement, respectively. This significantly lower
MDD, typical of a lightweight fill material,decreases the overburden on the foundation, which
is an advantage over traditional pavement subbase
materials.
Generally, bearing capacity as measured by CBRand swelling of the compacted materials arecontrolled by the fine content. Fig. 5 shows thesoaked CBR values at different compaction energy
levels in logarithm function. For a particular MD
replacement ratio, the soaked CBR significantly
increases with increasing compaction energy, E.
There are two linear slopes and the slope change isfound at E = 2681 kJ/m3 (modified Proctor energy).
The second slope (E > 2681 kJ/m3) is steeper than
the first one (E < 2681 kJ/m3), indicating that thecompaction energy is more significant on theincrease in soaked CBR when E > 2681 kJ/m3.

MELAMINE DEBRIS REPLACEMENT
The lateritic soil studied is composed of 21.7%
fine-grained particles and 78.3% coarse-grained
particles in which 25.6% is gravel and 31% is sand.
The specific gravity of coarse-grained particles is
2.67 and the liquid and plastic limits are 40.7% and20.9%, respectively. This compacted lateritic soil
has liquid limit (w L ) = 40.7%, plasticity index (I p ) =19.7%, California Bearing Ratio (CBR) = 9.3,swelling percentage = 6.4), which does not meet therequirement of the Department of Highways,
Thailand (required CBR are greater than 25% forsubbase), and will require blending with a higher
quality material if it is to be considered as apavement subbase or fill material.
Fig. 1 and Fig. 3 show the particle size
distribution curves of the lateritic soil/MD blends at
MD replacement ratios of 50%, and 10%,respectively compared with the upper and lower
boundaries of subbase materials specified by the
Department of Highways, Thailand. It is noted that
the MD particles larger than 3.80 mm (D 60 ) areabove the upper boundary and MD particles finer
than 1.05 mm (D 20 ) are below the lower boundary.
The results indicate that MD replacement reducesfine content of the lateritic soil. It is noted that the3

GEOMATE- Bangkok, Nov. 14-16, 2016100

5

Swelling (%)

80

Percentage passing (%)

6Upper boundLateritic soil Before compaction
Lateritic soil After compaction
Melamine Debris(MD) Before compaction
Melamine Debris(MD) After compaction
Lateritic soil : MD(50:50) Before compaction
Lateritic soil : MD(50:50) After compaction
Lower bound

60

40

20

Lat100Lat90:MD10
Lat80:MD20
Lat70:MD30
Lat60:MD40
Lat50:MD50

4
3
2
1
0

0
102

101

100

10-1

1000

10-2

Particle size (mm)

4000

500

E (kJ/m3)

Fig. 1 Particle size distribution of lateritic soil/MD
blend at 50% MD replacement.

Fig. 2 Swelling versus compaction
relationship for lateritic soil/MD blends.

100

energyreplacement ratios.
32

Upper boundLateritic soil Before compaction
Lateritic soil After compaction
Melamine Debris(MD) Before compaction
Melamine Debris(MD) After compaction
Lateritic soil : MD(90:10) Before compaction
Lateritic soil : MD(90:10) After compaction
Lower bound

80

28

60

24

CBR (%)

Percentage passing (%)

3000

2000

40

Lat50:MD50
Lat60:MD40
Lat70:MD30
Lat80:MD20
Lat90:MD10
Lat100

20
16

20

12
0
102

101

100

10-1

8

10-2

1000

Particle size (mm)

2000

3000

4000

500

3

Fig. 3 Particle size distribution of lateritic soil/MD
blend at 10% MD replacement.

E (kJ/m )Fig. 5 Soaked CBR versus compaction energy
relationship for lateritic soil/MD blends.

25

Dry Density (kN/m3)

Zero-air-void20

curve(Lateritic soilFig. 2 shows the swelling versus logarithm of E
relationship for various MD replacement ratios.
Similar to the soaked CBR versus logarithm of E
relationship, the E = 2681 kJ/m3 is regarded as thethreshold limit separating the first and second slopesof swelling versus logarithm of E relationship forlateritic soil.
Without MD replacement, E can reduce theswelling, especially when E > 2681 kJ/m3 as seenthat the second slope is remarkably lower than thefirst slope. The MD replacement significantly
reduces the swelling; i.e., the swelling of lateritic
soil reduces more than twice when only 10% MD is
replaced. The first and second slopes of the blends
are essentially the same and can be represented by agentle linear function when MD replacement isgreater 20%. It is noted that the swelling of theblends is significantly reduced even at a small E
level unlike the marginal lateritic soil (without MD),
which requires high E level to increase attractive
forces among soil particles.
Based on an analysis of soaked CBR andswelling test results, it is practical to relate thesoaked CBR and swelling of blends at various

Gs=2.75)

15
Lateritic Soil
Melamine Debris(MD)Lateritic Soil : MD(50:50)
Lateritic Soil : MD(60:40)
Lateritic Soil : MD(70:30)
Lateritic Soil : MD(80:20)
Lateritic Soil : MD(90:10)

10

5
4

6

8

10

12

14

16

18

Water content (%)

Fig. 4 Dry density versus moisture content
relationship for various lateritic soil/MD blends.
For a particular energy, the soaked CBR
increases with the MD replacement ratio; i.e., the
50% MD replacement exhibits the highest soaked
CBR for all E values. It is obvious that the first slopeof all the blends is essentially the same and thesecond slope of 50% replacement blend is thehighest. This implies that the rate of soaked CBRdevelopment with MD replacement ratio at E < 2681
kJ/m3 is essentially the same even with different MD
4

GEOMATE- Bangkok, Nov. 14-16, 2016

compaction energies in term of MD replacement. As
such, the predictive equations (Fig. 6 and Fig. 7) forsoaked CBR and swelling in term of MD
replacement ratio are presented as follows:
CBRMD
1.365exp 0.010 MD
CBR0
for 1197 kJ/m3 < E < 2681 kJ/m3
CBRMD
1.231exp 0.009 MD
CBR0
for 2681 kJ/m3
< E <
(2)

The formulation of the predictive equation is on
sound principle and can be fundamental to othermarginal soils.
FA GEOPOLYMER STABILIZATION
The marginal lateritic soil was collected from aborrow pit in Rayong province, Thailand. The
specific gravity is 2.58. The liquid limit, plastic limit,
and plastic index are 27.72%, 21.65%, and 6.07%,
respectively. This soil is classified as silty clayey
sand (SC-SM) according to the Unified Soil
Classification System and A-2-4(0) according to the
AASHTO system. The compaction characteristics
under modified Proctor energy are OMC of 8.0%,
and MDD of 20.85 kN/m3. CBR value is 14.7% at
95% of MDD. Los Angles (LA) abrasion is 52.9%.
With low CBR and high LA abrasion compared tothe specification of the Department of Highways,
Thailand (required CBR and LA abrasion are greaterthan 25% and less than 60% for subbase materials,
located below the base course, respectively), thislateritic soil is classified as a marginal soil. Fly Ash(FA) used in this study was obtained from Mae Moh
power plant in Northern Thailand, which is thelargest lignite power plant in Thailand.

(1)

3591

S MD
10.867 exp 0.06 MD
S0
for 2681 kJ/m3 < E < 3591 kJ/m3

kJ/m3

(3)

where CBR MD and CBR 0 are the soaked CBR at
different MD replacement ratios (ranging from 0%
to 50%) and soaked CBR at 0% MD replacement,
respectively and S MD and S 0 are the swelling atdifferent MD replacement ratios (ranging from 0%
to 50%) and swelling at 0% MD replacement,
respectively. These predictive equations are usefulfor predicting soaked CBR and swelling at different
MD replacement ratios based on the values oflateritic soil (without MD replacement). The
coefficients of correlation of these equations are
greater than 0.91, confirming the validity of theseequations.
3

CBRMD/ CBR0

CBRMD/CBR0 = 1.365exp(0.010MD)
2
R = 0.972
2

1

CBRMD/CBR0 = 1.231exp(0.009MD)2
R = 0.905

3

E1=1,197.03 kJ/m
E2=2,681.35 kJ/m 33
E3=3,591.09 kJ/m

0
0

10

30

20

40

50

MD Replacement (%)

Fig. 6 Relationship between normalized soaked CBR and MD replacement ratio.

560

GEOMATE- Bangkok, Nov. 14-16, 2016

Normalized Swelling, SMD/S00.7
0.6
SMD/S0 = 0.867exp(-0.06MD)
2
R = 0.98

0.5
0.4

E1=1,197.03 kJ/m 3
E2=2,681.35 kJ/m 3
E3=3,591.09 kJ/m 3

0.3
0.2
0.1
0
0

10

20

30

40

50

60

MD Replacement (%)

Fig. 7 Relationship between normalized swelling and MD replacement ratio.
kPa. This moderately high UCS at Na 2 SiO 3 :NaOH
The major components are 36.00% SiO 2 ,26.73% CaO, 17.64% Fe 2 O 3 , and 16.80% Al 2 O 3 .
ratio of 100:0 is contributed from the C-S-H, which
According to ASTM C 618-12, it is categorized asis the cementitious product from the reaction
class C high-calcium fly ash (CaO > 10%).
between silica from Na 2 SiO 3 and calcium from FA.
In presence of Ca2+, Na 2 SiO 3 forms soluble calcium
The liquid alkali activator was a mixture ofsilicate, which polymerizes further to form gels that
sodium silicate (Na 2 SiO 3 ) solution, composed of
15.50% Na 2 O, 32.75% SiO 2 , and 51.75% water bybind soil particles together and fill voids as theweight, and sodium hydroxide (NaOH) solution withfollowing equation [16]:5 molars.
Ca 2+ +Na 2SiO3 +mH 2 O o CaO  3SiO 2 .mH 2 O  2Na 
Fig. 8 shows the relationships between dry unit
weight and liquid alkali activator content with
(4)various Na 2 SiO 3 :NaOH ratios of lateritic soil-FA
geopolymer. The compaction curves of lateritic soilThe effect of C-S-H is minimal when the curing
FA geopolymer are different and depend on thetime is greater than 28 days as seen that the UCS ofratio.
Foraparticular
Na 2 SiO 3 :NaOH
sample at Na 2 SiO 3 :NaOH ratio of 100:0 isNa 2 SiO 3 :NaOH ratio, the dry unit weight of lateritic
essentially constant when the curing time is greatersoil-FA geopolymer increases with increasing liquidthan 28 days.
alkali activator content until MDD is attained at an
Since the N-A-S-H products are time-dependent,
optimum alkali activator content. Beyond thisthe UCS of geopolymer samples with the presence
optimum value, the unit weight decreases as theof NaOH (Na 2 SiO 3 :NaOH ratios of 90:10, 80:20,
alkali activator content increases. This characteristic
and 50:50) increases with curing time even after 28
is similar to a typical compaction behavior ofdays of curing. Due to the coexistence of C-S-H andcoarse-grainedmaterials.
The
optimum
N-A-S-H products, the geopolymer samples withNa 2 SiO 3 :NaOH ratio exhibiting the highest DMM
Na 2 SiO 3 :NaOH ratios of 90:10 and 80:20 exhibit
was found at 90:10.
higher 7-day UCS than the geopolymer sample withNa 2 SiO 3 :NaOH ratios of 100:0 (its UCS is mainly
Unconfined compressive strength
contributed from C-S-H). The optimum coexistenceof C-S-H and N-A-S-H products at 7 days of curing
The UCS of geopolymers is attributed to the(providing the highest UCS) is found atreaction between silica and alumina in the presence
Na 2 SiO 3 :NaOH ratio of 90:10.
of alkali ions and in age of specimens [13]. The UCS
Based on the strength data, it is suggested thatvalues of the lateritic soil-FA geopolymer for
the Na 2 SiO 3 :NaOH ratio of 50:50 is the best
various Na 2 SiO 3 :NaOH ratios (100:0 to 50:50) andingredient whose UCS is greater than 2,413 kPa
curing times (7 to 90 days) are presented in Fig. 9. It(required for heavy traffic). The highest rate of UCS
is evident that the UCS values for all
development of the geopolymer at this ingredient is
Na 2 SiO 3 :NaOH ratios increase as the curing time
also observed.
increases. At early stage of geopolymerization, theMicrostructural analysis
maximum 7-day UCS is found at Na 2 SiO 3 :NaOH
ratio of 90:10 (about 7,100 kPa), which is greater
Fig. 10 shows the microstructure of the fourthan the strength requirement specified by theNa
2 SiO 3 :NaOH ratios of 100:0, 90:10, 80:20, andnational road authority in Thailand (UCS > 1,724
50:50
at 7 days of curing (early stage ofkPa for light traffic and UCS > 2,413 kPa for highgeopolymerization process) at ambient temperature.
traffic) [14-15]. The 7-day UCS at Na 2 SiO 3 :NaOH
ratio of 100:0 (no NaOH) is approximately 5,800
6

Na2SiO 3:NaOH

22
Dry unit weight (kN/m3)

Unconfined compressive strength (kPa)GEOMATE- Bangkok, Nov. 14-16, 2016

100:0
90:10

80:20
50:50

20

18

16
10

15

20

Ambient temperature

20000
Na2SiO 3:NaOH
17500 100:0
90:10
15000

80:20
50:50

12500
10000
7500
5000
2500

Heavy traffic (2,413 kPa)

00

Light traffic (1,724 kPa)

10

25

20

30

40

50

60

70

80

Curing time (days)

Fig. 9 Effect of curing time on UCS of lateritic soil–FA
geopolymer

Molding moisture content (%)

Fig. 8 Compaction curves of lateritic soil – FA
geopolymer at different ingredients.

(a)

(b)

(d)
(c)
Fig. 10 SEM images of lateritic soil – FA geopolymer cured at 7 days of curing at ambient temperature forvarious Na 2 SiO 3 :NaOH ratios of (a) 100:0, (b) 90:0, (c) 80:20, and (d) 50:50.

(a)

(b)

7

90

GEOMATE- Bangkok, Nov. 14-16, 2016(d)(c)
Fig. 11 SEM images of lateritic soil – FA geopolymer cured at 90 days at ambient temperature for various
ingredients of Na 2 SiO 3 : NaOH (a) 100:0, (b) 90:10, (c) 80:20, and (d) 50:50.
The cementitious (C-S-H and N-A-S-H) productson the high calcium FA surface are clearly observedat Na 2 SiO 3 :NaOH ratios of 90:10 and 80:20 whencompared to those at Na 2 SiO 3 :NaOH ratios of 50:50
and 100:0, respectively Many holes on the FA
surface are clearly observed for Na 2 SiO 3 :NaOH
ratios of 50:50 due to strong alkaline reaction fromNaOH while the least etched holes on FA surface areobserved at Na 2 SiO 3 :NaOH ratio of 100:0.
The effect of Na 2 SiO 3 :NaOH ratio on the growthof geopolymerization products is illustrated bycomparing Fig. 11a–11d, which show SEM imagesof the lateritic soil–FA geopolymer samples atvarious Na 2 SiO 3 :NaOH ratios of 100:0, 90:10,
80:20, and 50:50 after 90 days of curing. For
Na 2 SiO 3 :NaOH ratio of 100:0 (no NaOH), the FA
surface is approximately smooth and spherical withtraces of C-S-H products. This indicates that thesilica and alumina in FA was insignificantly leachedout even after a very long curing time and the UCSof the samples is contributed from C-S-H (reactionof Na 2 SiO 3 and CaO from FA) (Fig. 11a).
The NaOH content is not sufficient for leachingsilica and alumina from FA for the sample at
Na 2 SiO 3 :NaOH ratio of 90:10 as seen by the small
amount of cementitious products on FA surface (Fig.
11b) when compared to the samples at
Na 2 SiO 3 :NaOH ratios of 80:20 and 50:50 (Fig. 7cand d). For both samples at Na 2 SiO 3 :NaOH ratios of
80:20 and 50:50, the etched holes are mostly filledwith other smaller size ash particles and thecementitious products fill up the pore space,resulting in a dense matrix. The cementitious
products are more for the sample at Na 2 SiO 3 :NaOH
ratio of 50:50. As such, the maximum UCS is foundat Na 2 SiO 3 :NaOH ratio of 50:50 at 60 and 90 daysof curing (vide Fig. 9)
The present study on the improvement ofmarginal lateritic soil by FA based geopolymer has
significant impacts on sustainable pavement
applications. The green geopolymer stabilizedlateritic soil is proved as suitable where the 7-day
UCS meets the specification for heavy traffic andthe UCS increases with increasing curing time evenafter 28 days of curing due to the growth ofgeopolymerization product (N-A-S-H) over time.
CONCLUSION
This paper investigates physical andmechanical properties of marginal lateritic soil
blended with MD at various replacement ratios toascertain it as an engineering fill material as well as
UCS and microstructural characteristics of themarginal lateritic soil stabilized with high calcium
FA based geopolymer as low-carbon alternative
green bound pavement material.
The marginal lateritic soil improvement by MD
replacement is found to be a sustainable engineering
fill material based on this research. The laboratory
evaluation of the lateritic soil/MD blends at different
MD replacement ratios included physical property
tests such as particle size distribution, Atterberg
limits and LA abrasion, as well as mechanicalproperty tests such as soaked CBR and swelling.
MD replacement can improve the mechanicalproperty requirement for engineering fill material
according to national local authority. The MD
traditionally destined for landfill can be used as areplacement material to stabilize lateritic soil fordeveloping the sustainable engineering fill material.
With 20% MD replacement, the physical andmechanical properties of blends meet therequirement. The 50% MD replacement blend is at
the borderline of the specification for subbase
material.
It is evident from this study that the strength oflateritic soil-FA geopolymer meets the standard
requirements for heavy and light traffic bound basematerials specified by Department of Highways(DOH) and Department of Rural Roads (DRR),
Thailand. FA based geopolymer can be used as analternative binder to PC for sustainable pavement
applications. This study indicates that the marginal
lateritic soil can be stabilized by the high calcium
FA based geopolymer to be a bound pavement
material for both light and heavy traffic. This willleads to a usage reduction of PC and develop a
possible environmentally sustainable material. The

8

GEOMATE- Bangkok, Nov. 14-16, 2016

economical ingredient is
Na 2 SiO 3 :NaOH of 50:50.

suggestedtobeMaterials in Civil Engineering”, Vol. 24(8),2012 pp. 1059-1066.
[8] Kampala A, Horpibulsuk S, Prongmanee N, and

ACKNOWLEDGEMENTS

Chinkulkijniwat A, “Influence of Wet-Dry
Cycles on Compressive Strength of Calcium
Carbide Residue–Fly Ash Stabilized Clay”,
Journal of Materials in Civil Engineering, Vol.
26(4), 2014, pp. 633-643.
[9] Puppala AJ, Wattanasanticharoen E, andPunthutaecha K, “Experimental evaluations ofstabilisation
methodsforsulphate-rich
expansive soils”, Ground Improvement, 7(1),2003, pp. 25-35.
[10] Sukmak P, De Silva P, Horpibulsuk S, andChindaprasirt P, “Sulfate resistance of clayPortland cement and clay high-calcium fly ash
Geopolymer”, Journal of Materials in Civil
Engineering, Vol. 27(5), 2015, pp.04014158(111).
[11] DH-S, Standard for Highway Construction. DHS205/2532, 1996, Bangkok.
[12] Horpibulsuk S, Katkan W, Naramitkornburee A,“Modified Ohio’s curves: A rapid estimation ofcompaction curve for coarse-and fine-grainedsoils”, Geotechnical Testing Journal, Vol. 32(1),2009, pp.64-75.
[13] Somna K, Jaturapitakkul C, Kajitvichyanukul P,
and Chindaprasirt P, “NaOH-activated ground
fly ash geopolymer cured at ambient
temperature”, Fuel, Vol. 90(6), 2011, pp. 21182124.
[14] DOH, DH-S204/2000 Standard of Soil Cement
Base: Department of Highways, 2000.
[15] DRR, DRR244-2013 Standard of Soil Cement
Base: Department of Rural Roads, 2013.
[16] Brykov AS, Danilov VV, Korneev VI, andLarichkov AV, Effect of hydrated sodium
silicates on cement paste hardening. Russian
Journal of Applied Chemistry, Vol. 75(10),2002, pp. 1577-1579.

This work was supported by the Thailand
Research Fund under the TRF Senior Research
Scholar program Grant No. RTA5680002 andSuranaree University of Technology.
REFERENCES[1] Horpibulsuk S, Katkan W, Sirilerdwattana W,
Rachan R, “Strength development in cement
stabilized low plasticity and coarse grainedsoils: Laboratory and field study”, Soils andFoundations, Vol. 46(3), 2006, pp. 351-366.
[2] Horpibulsuk S, Katkan W, Apichatvullop A,“An approach for assessment of compaction
curves of fine-grained soils at various energies
using a one point test”, Soils and Foundations,Vol. 48(1), 2008, 115-125.
[3] Arulrajah A, Ali MMY, Disfani M.M. andHorpibulsuk S, “Recycled glass blends inpavement base/subbase applications: laboratory
and field evaluation”, Journal of Materials in
Civil
Engineering,
2014,
doi:
10.1061/(ASCE)MT.1943-5533.0000966.
[4] Siwadamrongpong S, Wonglane K, and Mahai
M, “Influence of water absorption themechanical properties of recycled materials
from waste melamine”, South East AsianTechnical University Consortium. 6 - 7 March
2012. KMUTT. Bangkok, Thailand.
[5] Chinkulkijniwat A, and Horpibulsuk S, (2012).
“Field strength development of repaired
pavement using the recycling technique”,
Quarterly Journal of Engineering Geology andHydrogeology, Vol. 45(2), pp. 221-229.
[6] Davidovits J, Geopolymer Cement. Geopolymer
Cement, a review, 2013.
[7] Ebrahimi A, Edil TB, and Son YH,“Effectiveness of Cement Kiln Dust in
Stabilizing Recycled Base Materials. Journal of9Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ESTIMATION OF DYNAMIC BUCKLING STRENGTH OF CIRCULAR
TUBE PILES WITH LIQUEFIED SOIL
Moeko Matoba1, Tenshiro Goto1 and Yoshihiro Kimura1
1
Tohoku University, Japan

ABSTRACTIn current design criteria, the possibility of dynamic buckling of a pile foundation is not considered. Designcodes have no prescription for limitations of a slenderness ratio for piles. However, pile buckling occurred whenslender steel piles beneath buildings experience high axial compression forces as a result of vertical loads
increased by the over turning moment of buildings, generated by inertia forces of buildings. With coincidence ofliquefaction, the buckling risk is most likely to be even higher. In this study, centrifuge tests are conducted toassess the dynamic buckling behavior of circular tube piles with liquefied soil. Test parameters were the natural
period of a superstructure–foundation system, input wave, pile slenderness ratio, and the material properties ofthe pile. Dynamic buckling stress measured using a centrifuge test can be evaluated approximately with anequivalent slenderness ratio of elastic buckling stress presented in the Limit State Design of Steel Structures.
Keywords: Steel pile, Dynamic buckling behavior, Liquefaction, Centrifuge tests

forces. The elastic flexural buckling equations ofpiles in liquefied soil were developed using theenergy method. These equations are applied to the
proposed equivalent slenderness ratio. Results showthat experimentally obtained values agree well withthe buckling curves having this slenderness ratio.

INTRODUCTION
Steel piles are believed to be able to sustain
strong lateral and compressive forces and that no
pile experiences flexural buckling because of thehigh compressive force attributable to theoverturning moment in addition to the dead load of
the building because of soil resistance against lateral
deformation of the piles. In fact, no damage of steel
piles, which have high strength and ductility, hasbeen identified in surveys of past earthquake damage.
However, during the 2011 Great East Japan
Earthquake, many piles of buildings in Sendai city,
sustained heavy damage because these buildings
were used even though their piles had alreadysustained heavy damage during the 1978 Miyagi
Earthquake [1]. It is difficult to assess damage topiles under the buildings in the soil if the piles can
carry a vertical load after the earthquake, even after
the piles have buckled during soil liquefaction. In
the Japanese current design code, the possibility ofsteel pile buckling is not considered. Design codes
have no prescription for limitations of theslenderness ratio for piles. The horizontal stiffnessof the ground decreases drastically whenliquefaction occurs during an earthquake. Moreover,when slender steel piles beneath buildings
experience high axial compression forces caused by
the overturning moment of the superstructure, which
are generated by inertia forces of buildings, flexural
buckling of piles occurs as presented in Fig. 1.
This study conducted centrifuge tests ofsuperstructure-pile-liquefied soil systems toinvestigate the dynamic buckling behavior of
circular tube piles subjected to compressive axial

Dead Load

Inertia Force

Horizontal Force due
to Inertia Force
Liquefied
Soil

Non
Liquefied Soil

Fig. 110

Force Couple due toOverturning Moment

Flexural Bucklingof Steel pile
Soil
Reaction

Concept for Flexural Buckling of Steel Pilein Liquefied Soil

GEOMATE- Bangkok, Nov. 14–16, 2016PP
Elastic
Springcr

Soil Reaction
Distribution

Elastic Eigen Value
Eqs. (4) and (5)Smaller value from Eqs. (4), (5) and (6)(kN/mm2)
3l

Kc=1MN/m2

2

Kc=0.1MN/m2
Kc=0.01MN/m2

1

Fig. 2Euler Buckling

Analytical Model of Steel Pile in Liquefied
Soil and Distribution of Subgrade Reaction.

00ELASTIC BUCKLING LOAD FOR A STEELPILE IN LIQUEFIED SOIL

Fig. 3For this chapter, to develop the elastic bucklingload of a steel pile in liquefied soil, the following areassumed. The external force subjected to a pile isonly a vertical load such as Fig. 2. Both ends of thepile are fixed. The subgrade reaction is replaced byelastic horizontal springs.
When a steel pile in liquefied soil is buckled
laterally, the potential energy is expressed as:U

1 l EIu 2  Pu 2  Ku 2 dx
2 0

2Kc
(l  x)l

4(5)

The validity of the pile buckling load calculated
from Eqs. (4) and (5) is ascertained by elastic
eigenvalue analysis, where the pile is replaced by aline element divided into 50 parts in the longitudinal
direction. The coefficient of the subgrade reaction is
replaced by horizontal springs.
Figure 3 presents the relation between the elastic
buckling stress for steel piles with the subgrade
reaction and slenderness ratio. The plots represent
elastic eigenvalue analysis results. The thin broken
curves are drawn from Eqs. (4) and (5). The triangle
subgrade reaction distribution is applied with thecoefficient of the subgrade reaction, such as 1
MN/m2, 0.1 MN/m2, and 0.01 MN/m2. When theslenderness ratio and the coefficient of the subgrade
reaction are small, the pile buckling stresses
calculated from Eqs. (4) and (5) are almost equal tonumerical analysis results. However, when theslenderness ratio and the coefficient of the subgrade
reaction are large, the pile buckling stresses
calculated from Eqs. (4) and (5) are higher than
numerical analysis results (e.g. Kc is 1 MN/m2).
For a liquefaction layer thicker than the pilelength, taking a local minimum of the buckling load,
the closed form solution shown below is suggested.

(1)

(2)

(3)

By substituting Eq. (3) into Eq. (2) and partiallydifferentiating Eq. (2) the buckling load is obtained.
 
Pcr    EI l 

Analytical Model of Steel Pile in Liquefied
Soil and Distribution of Subgrade Reaction.
3 l  K

A triangular distribution is applied to the
distribution of the subgrade reaction as in Fig. 2 andthe following.

K

150 =(l/i)

100  4   c
4    EI

where EI stands for the steel pile flexural rigidity, lsignifies the liquefaction layer thickness, and K
denotes the subgrade reaction coefficient. Also, P
and u respectively represent the pile vertical load
and lateral displacement. The steel pile lateral
deformation is expressed as a sine curve function.
2 x u  a  1  cos

l 


50

 

  8 1 
 

2

1 
l 
  2 
3
 

2

Kc   3 1      
EI   4 2  l 

l  2 EI / 3K  (6)
4

(4)

11

EI 

K c 

2c

GEOMATE- Bangkok, Nov. 14–16, 2016

p/p1
3

110
p1=kh1By

bending plate X

2

y/y51
pile B，D

200

Thick solid curves from Eqs. (4), (5),and (6) fit the numerical analysis results
well. The subgrade reaction coefficientis adopted as the following in [2].
2

3

4

kh1  80 E0 B (MN/m )80
superstructure

60
pile D

pile C

(8)

40
80

3

restrictingtool

In those equations, kh1 consists of thedeformation modulus and the pile
diameter.
Figure 4 shows the relation between
the subgrade reaction and lateraldisplacement of the soil. The solid curveis the subgrade reaction curve as
described in [2]. The broken line passes
through the origin and point (1.0,1.0) of
the coordinates has the gradient defined as Kc in Eq.
(7). As described in this paper, soil having this curveor this line is designated as “elasto-plastic soil” or“elastic soil.” The secant rigidity decreases as thelateral displacement of soil increases in Fig. 4. Whenthe lateral displacement at the buckling of the piles
is greater than reference displacement y/y1=1, thecoefficient of the subgrade reaction in elasto-plastic
soil is smaller than that in elastic soil. Therefore, thecoefficient of the subgrade reaction in elasto-plastic
soil should be revalued when assessing lateral soil
displacement.
As described in this paper, the lateraldisplacement at the buckling of the piles is assumed
to be 0.002 l in [3]; this secant rigidity is designatedas Kc’. The reduction ratio of the soil subgrade
reaction,  is approximated as the ratio of Kc’ to Kc.

  l0 / l

A7(c) Enlarged View
of Pile A(a) Side(7)

K c =kh1 B (MN/m )A6

unit:mmlaminar box

A5

150

Subgrade Reaction Curve.

280

4

60

3

A4

60

2

pile A，C

A3

75

1

100

yi

260

0

elastic soilxi

40 40 40 40 40 40

restricting
tool(DBF, AIJ )[2]

Kc

A1
A2

0

kh=kh1(y/y1)-0.5

0laminar box

hY

pile cap

1Fig. 460

superstructure
elasto-plastic soil

strain gauge
accelerometerwater pressuregauge

pile B

pile A

450(b) Top
Fig. 5Specimen and Instrumentation.

when the piles buckle. Therefore, 0.1 or 0.2 for soil
relative density of 30% or 60% is adopted as the
reduction rate because of soil liquefaction [2].
CENTRIFUGE TESTSAs described in this chapter, centrifuge tests ofthe superstructure–pile foundation model wereconducted to clarify dynamic buckling behaviors ofcircular tube piles.
Outline of Centrifuge Test

Figures 5(a) and 5(b) show the specimen and the
instruments. The specimen has a superstructure withmass, bending plates, four piles, and a saturated sandlayer. The width of the foundation andsuperstructure is 80 mm. The superstructure heightis 60 mm. The pile cap is laterally fixed to subject
the piles to varying axial force caused by thesuperstructure weight and the overturning momentof the superstructure. As presented in Fig. 5, strain
gauges on the left surface of piles and the bending
plate are represented xi, the gauges on the right(9)

Therein, l denotes the pile length, l0 is 5 m. By
substituting Kc’ into Kc in Eqs. (4)–(6), the pilebuckling stresses in elasto-plastic soil are calculated.
Furthermore, for centrifuge tests, the coefficientof the subgrade reaction decreases in liquefied soil

12

GEOMATE- Bangkok, Nov. 14–16, 2016

surface are represented as yi. Here, i represents 1–5
of each pile height, and 1–3 for each bending plateheight. Accelerometers are put on upper surface ofthe superstructure and the pile foundation, the plateat bottom of the pile. Furthermore, accelerometers
and water pressure gauges are placed at 80 mm, 140mm, and 200 mm from ground level as presented inFig. 5(a). The boundary condition of the piles isperfectly fixed. The pile cap is rotationally free, butlaterally fixed. Centrifuge tests were performed
under centrifugal acceleration of 40 g with acentrifugal loading device at the Disaster Prevention
Research Institute, Kyoto University.
Table 1 presents the specimen specifications.
The pile material is aluminum A1050S and brass
C2680, the bending plate material is A5052; theothers are SS400.
Figure 6 presents material properties of the piles.
Therein, E is Young's modulus, y is the yielding

Table 1 Specification of the Specimens
Superstructure: m1
Bending plate
Pile cap: m2

Pile

Weight
1.95×106(N)
Thickness
80(mm)
Weight8.3×105(N)
Diameter
240(mm)
Thickness
20(mm)

stress, and Est is the strain hardening gradient.
Table 2 shows specimen parameters, which are
the pile length, the initial axial force ratio, thenatural period of the superstructure, the relative
density, and the maximum input wave. Case 1 series
and Case 2 series were conducted, respectively,
without soil and with liquefied soil. The initial axial
load ratio of aluminum piles, N0/Ny is 0.37, issued
from the design axial force of steel piles in actual
structures. The initial axial load ratio of brass piles,
N0/Ny, is 0.13, to apply the same mass for specimens
with aluminum piles. The pile length of the Case 1
series is 6000 mm or 9600 mm in full-scale. The pilelength of Case 2 is 10400 mm. The natural period of
the superstructure is presented in Table 2.
Figures 7(a) and 7(b) present the time history of
the input waves. The sweep wave portrayed in Fig.
7(a) changes the period from 2.0 to 0.3 s in 50 s. The
maximum input acceleration of the sweep wave is
1.0–5.5 m/s2. The maximum input acceleration of(N/mm )(N/mm )

2

2

150
100
50

u(N/mm2)145

y(N/mm2)136

l(N/mm2)

115

400
200

(N/mm ) 6.85×10
st(N/mm2) 727
2

0 E
0 0.2 0.5

14

1.5(%)(a) Aluminum
Fig. 6u(N/mm2)498

y(N/mm2)398

l(N/mm2)275

(N/mm ) 1.03×105
st(N/mm2) 1.72×104
2

0 E
0 0.2

1

2 (%)(b) BrassMaterial Properties of the Piles.

Table 2 Specimen Parameters
Relative
Pile
InitialPlate
Maximum Input State ofperiod of superDensity Input Wave
Specimen Material length
axial
lengthWave (m/s2) vibration
structure (s)Dr (%)l (mm) force ratio h (mm)1.0
Elastic
Case 1-1
2.5
Collapse
1800
0.79
Case 1-2
4.5
Collapse
6000
1.0
Elastic
Case 1-3 aluminum
0.37
2.5
Collapse
1400
0.59
Sweep
Case 1-4
3.0
Collapse
1.0
Elastic
Case 1-5
1800
0.79
1.8
Collapse
9600
1.0
Elastic
Case 1-6 brass
0.13
1400
0.59
5.5
Collapse
Case 2-1
1400
0.59
30
Collapse
Case 2-2
1800
0.79
30
Collapse
Sweep
3.0
aluminum
0.37
Case 2-3
60
Collapse
10400
Case 2-430
Coastal Wave
6.9
Collapse
1400
0.59
Case 2-5 brass
0.13
30
Sweep
3.0
Plasticized※Full-scale

13

GEOMATE- Bangkok, Nov. 14–16, 2016

2

2

2

a (m/s )

10
0

0
-2
0

10

20
30
40
(a) Sweep Wave

-10
0

sec

50

10

20

30

sec

Fig. 10 Acceleration Response Time story of theSuperstructure.

210

a (m/s )a (m/s )30

N (×10 kN)
b
2.0

-10
0

10

Fig. 7203040(b) Coastal Wave

sec

50

0

Ny=yA△-2.0
0

Input Wave Time Histories.

10

20

30

sec

Fig. 11 Axial Force Response Time History.

2r (m/s )ah=0.02 (%)

sweep wave
coastal wave

4

-1

0.59

0

Fig. 8
G

0.5

0.79

1

1.5

pileB,C,D

▼

0

2

0

pileA

1 bi







sec

Fig. 12 Bending Strain Response Time History.

T(s)

1

Response spectrum of the input wave.

 (%)

pileA

ci

pileB,C,D▼

0
u

Case1-1
Case1-3
Case2-1

6

-1


4

Fig. 9





sec

Fig. 13 Axial Strain Response Time History.

2
0
0



Dynamic Buckling Behavior of Superstructure–
Pile Foundation System
0.5

1Transfer Function
Superstructure.

1.5

of

pile

T(s)

cap

For superstructure–pile models, this section
clarifies the dynamic buckling behavior of circulartube piles.

andTime history

the coastal wave is 6.0 m/s2.
Figure 8 portrays the acceleration response
spectrum of the input wave. Figure 9 shows atransfer function of pile foundation andsuperstructure. The damping ratio is 0.02. The
natural periods of Case 1-1 and Case 1-3 shown in
Table 2 are 0.59 s and 0.79 s, which are close to theperiod at the maximum value of the coastal wave
spectrum.

Figures 10–13 present time histories of Case 1-1
shown in Table 2.
Figure 10 shows the acceleration response time
history of the superstructure. Acceleration gradually
increases. It reaches the maximum value at 19 s
(1.06 Hz). Here, 1.06 Hz is the frequency of thesweep wave at dynamic buckling. Then theacceleration response fluctuates suddenly because of
the pile collapse.
Figure 11 shows the axial force of pile responsetime history. The pile axial force is calculated as:

14

GEOMATE- Bangkok, Nov. 14–16, 2016Nb 

 bxi   byi
2

EAb / 2  m2 g / 4(10)3N (×10 kN)bwhere it is assumed that two piles uniformly carryvarying axial force of the bending plate. Here, Ab
denotes the bending plate cross-section bxi ， byi
represents the bending plate strain, and m2 signifies
the pile cap weight. Results show that the bending
plates for all specimens remain elastic until the piles
collapse. The horizontal broken line in Fig. 11
shows the piles’ yield axial load. The white triangle
symbol in Fig. 11 represents the maximum
compression axial force before pile collapse. The
maximum axial force is less than the yield axial
force issued from dynamic buckling.
Figures 12 and 13 respectively depict bendingstrain and axial strain response time histories. These
values are calculated with strain gauges at the pile
center, as shown in Fig. 5. The bending strain andaxial strain are calculated using the following
equations.

Case1-1

0

-1.0

N
-2.0
-0.8

y

△

0 ci (%)

-0.4

Fig. 14 Relation between Axial Force and Axial
Strain3〇

N (×10 kN) △b

Elastic gradientinitial axial force
maximum axial force

Case1-1

 bi 
 ci 

 pxi   pyi
2
 pxi   pyi
2

0

(11)
(12)

-1.0

The red triangle symbols in Figs. 12 and 13
indicate the times for the maximum bending strain
increment. Black lines in Figs. 12 and 13 are theresponse time history of the pile that starts dynamic
buckling first. Gray lines show the other piles’
respective response time histories.
For Case 1-1, pile A buckled first. When pile A
carries a maximum axial compressive force, thebending strain increment also reaches the maximum
value. The bending strain of pile C increases
immediately after the maximum bending strain
increment of pile A. The bending strain of piles C
and D is increased suddenly a few seconds later thanthat of pile A. Therefore, it can be inferred that pile
C buckled immediately after dynamic buckling ofpile A, and that piles B and D buckled dynamically.
Results show that the maximum axial strain isreached after the maximum bending strain increment
for all specimens.
Figure 14 shows the relation between axial force
and axial strain for the pile of Case 1-1. The verticalaxis is the pile’s axial force calculated using Eq. (10).
The horizontal axis is the axial strain at the center ofpiles calculated using Eq. (12). The black line
represents the hysteresis loop of the pile buckled
first. Gray lines represent the elastic gradient; Ny
represents the yield axial load of the piles. The circle
symbol in Fig. 14 shows the initial axial force. The
triangle symbol shows the maximum axial force.
Elastic hysteresis loops drawn with black lines arealmost identical to the elastic gradient drawn with

N
-2.0

△

y

-0.8

-0.4

0 bi (%)

Fig. 15 Relation between Axial Force and Bending
Straingray lines. Axial force for Case 1-1 with 76.8 ofslenderness ratio decreases gradually after the axial
force reaches the maximum value.
Figure 15 shows the relation between the axial
force and the pile bending strain. For Case 1-1, thebending strain and axial force increase gradually
until axial force reaches its maximum value. The
hysteresis loops depicted in Figs. 14 and 15 arealmost identical to the loops of the axial
compressive members with static buckling.
Pile dynamic buckling behavior
Figure 16 shows the bending strain distributionof piles that buckled first for Case 1-1, Case 1-3, andCase 1-5. Each plot presents the bending strain at themeasure point at 0.1% and 0.2% of the maximum
bending strain increment and the maximum axial
force. Maximum bending strain occurs at the pile
center for all specimens. Bending strain of the pile
end is less than that at the center of the pile.
Photo 1 shows the pile deformation of Case 1-6

15

GEOMATE- Bangkok, Nov. 14–16, 2016

bmax=0.1%

Maximum b incrementbmax=0.2%
l (m)Maximum axial force

l (m)

6

l (m)

6
8
4

4

(b) Center of pile
4
2

2

0
-0.4

0

0
0.4 -0.4(a) Case 1-1

00
0.4 -0.4(b) Case 1-30bi(%)

0.4(a) Whole piles(c) Case 1-5(c) End of Pile

Fig. 16 Bending Strain Distribution of Pile.

Photo 1 Deformation of Piles (Case 1-6).

after vibration. All piles deform laterally at thecenter, which is similar to bending strain distributionof piles such as Case 1-1 in Fig. 16. Local bucklingalso occurred at the pile end because of the pile’s
large flexural buckling in Photo 1(c).
Results show that the bending deformation of all
specimens appears to be almost identical.

s in Fig. 18. Furthermore, the bending strain
increases suddenly at about 19 s, as marked by the
red triangle symbol in Fig. 20, as soon as the axial
force reaches the maximum value shown by thewhite triangle symbol in Fig. 19. For Case 2-1, pile
B buckled first.
Figures 22–26 represent the time history of Case
2-2 presented in Table 2. Figure 22 shows theacceleration response time history of thesuperstructure. Acceleration increases gradually andreaches a maximum value at 13 s. Figure 23 depicts
the excess pore water pressure ratio time history.
Figure 24 portrays the axial force of the pile
response time history, which is calculated from dataof axial strain gauges on bending plates, as presented
in Fig. 5(a). Figures 25 and 26 respectively display
the bending strain and axial strain response time
history.
For Case 2-2, soil liquefaction occurs at about 14s, as presented in Fig. 23. Furthermore, then, thebending strain increases suddenly at about 15 s, as
denoted by the red triangle symbol in Fig. 25
immediately after the axial force reaches the
maximum value, as indicated by a white triangle
symbol in Fig. 24. For Case 2-2, pile A buckled first.
The response behaviors for Case 2-2 are similar to
those for Case 2-1, even though the natural period ofCase 2-2 is longer than that of Case 2-1.

Dynamic Buckling Behavior of Superstructure–
Pile–Liquefied Soil System

In this section, the dynamic buckling behavior of
circular tube piles is clarified with superstructure–pile–liquefied soil models.
Response time history
Figures 17–21 show the time history of Case 2-1
presented in Table 2. Figure 17 shows thesuperstructure acceleration response time history.
Acceleration increases gradually and reaches amaximum value at 19 s.
Figure 18 shows the excess pore water pressure
ratio time history. Black and gray curves
respectively show data from water pressure gauges
located on 80 mm and 140 mm from the groundsurface depicted in Fig. 5(a). The dotted line in Fig.
18 shows the time at which the excess pore waterpressure ratio reached 1.0. As described in this paper,
soil liquefaction occurs when the excess pore waterpressure ratio reaches 1.0.
Figure 19 portrays the axial force in the pile
response time history, as calculated using axial
strain gauges on bending plates (Fig. 5(a)). Figures
20 and 21 respectively show the bending strain andaxial strain response time histories.
For Case 2-1, soil liquefaction occurs at about 16

Pile dynamic buckling behavior
Figure 27 portrays the relation between theexcess pore water pressure ratio and the pile bendingstrain for Case 2-1. The vertical axis represents theexcess pore water pressure ratio. The horizontal axis
represents bending strain at the center of pile B, ascalculated from Eq. (12). Gray dotted and black

16

GEOMATE- Bangkok, Nov. 14–16, 20168a (m/s2)8

a (m/s2)

0

0
-8
0

5

10

15

20

25

-8
0

sec

0
0

water pressure gauge 1
water pressure gauge 2

5

10

15

20

25

10

15

20

25

sec

Fig. 22 Acceleration Response Time History of theSuperstructure (Case 2-2).

Fig. 17 Acceleration Response Time History of theSuperstructure (Case 2-1).
ru
1

5

water pressure gauge 1
water pressure gauge 2

ru
1

0
0

sec

5

10

15

20

25

sec

Fig. 18 Excess Pore Water Pressure Ratio Time
History (Case 2-1).

Fig. 23 Excess Pore Water Pressure Ratio Time
History (Case 2-2)

Nb (×103kN)
0

Nb (×103kN)
0

-2.0
0

Ny

△

5

10

15

20

25

sec

-2.0
0

Ny

△

5

10

15

20

25

Fig. 19 Axial Force Response Time History (Case
2-1).

Fig. 24 Axial Force Response Time History(Case 2-2).

b (%)

b (%)
0.8
0.4
0
-0.4


0.8
0.4
0
-0.4


pile B

pile A,C,D

▼











sec

 (%)

pile B

pile A,C,D

▼











sec

pile B,C,D

▼











sec

Fig. 25 Bending Strain Response Time History(Case 2-2).

Fig. 20 Bending Strain Response Time History(Case 2-1).
0.4 c
0
-0.4
-0.8


pile A

sec

 (%)

0.4 c
0
-0.4
-0.8


pile A

pile B,C,D

▼









Time

sec

Fig. 21 Axial Strain Response Time History (Case
2-1).

Fig. 26 Axial Strain(Case 2-2).

solid curves respectively show the hysteresis loops
before buckling occurs and after buckling occurs.
The bending strain increases only slightly before
the excess pore water pressure ratio reaches 1.0.
After liquefaction, the bending strain increases
suddenly because of the reduction of the horizontal
stiffness of the soil.

Figure 28 depicts the relation between the axial
force and pile bending strain for Case 2-1. The
vertical axis shows the axial force; the horizontal
axis represents bending strain at the center of pile B.
Gray dotted and black solid lines respectively show
hysteresis loops before liquefaction and afterliquefaction.

17

Response

HistoryGEOMATE- Bangkok, Nov. 14–16, 2016

increases only slightly, even if the axial force
increases. After liquefaction, the axial force reaches
a maximum value. Then dynamic buckling occurs.
The more bending strain increases after buckling,
the less axial force decreases. In contrast to thesuperstructure–pile models without soil in Fig. 15,
the axial force decreases only slightly until thebending strain reaches about 0.15%.
Photo 2 shows the pile buckling deformation for
the Case 2 series. For the specimens of Cases 2-1, 22 and 2-4, the piles’ flexural buckling occurs withlocal buckling at the pile end. For Case 2-3, with soil
relative density of 60%, pile buckling occurs
laterally near the pile head. In contrast, although pile
instantaneously becomes plastic for Case 2-5, which
has a smaller initial axial force ratio than the others,
the piles never buckle.

Before soil liquefaction, the bending strain
Before Buckling
After Buckling

ru

▼

1
0.8
0.6
0.4
0.2
Case2-10-0.4

bi (%)

0

DYNAMIC BUCKLING STRENGTHSTEEL PILES IN LIQUEFIED SOIL

Fig. 27 Relation between Excess Pore Water
Pressure Ratio and Pile’s Bending Strain.
3

N (×10 kN)b

OF

Figure 29 portrays the relation between buckling
stress and the equivalent slenderness ratio. The
vertical axis represents the value of the dynamic
buckling stress of the piles divided by the yielding
stress. The horizontal axis shows the equivalent
slenderness ratio of piles, calculated as:

Before Liquefaction
After Liquefaction0c  Py / Pcr

(13)

-1.0

where Py stands for the yield load and Pcr represents
Euler’s buckling load, as described in [4]. When
steel piles are braced continuously by the soil,
Euler’s buckling load is replaced by the bucklingload obtained from Eqs. (4)–(6). Then thisequivalent slenderness ratio is defined as themodified equivalent slenderness ratio.
Figure 29 presents the relation between theCase2-1

-2.0
-0.4

0

bi (%)

Fig. 28 Relation between Axial Force and Bending
Strain.

(a) Case 2-1(b) Case 2-2(c) Case 2-3(d) Case 2-4

Photo 2 Deformation of Superstructure – Pile Cap Liquefied Soil System.
18(e) Case 2-5

GEOMATE- Bangkok, Nov. 14–16, 2016

Case2-1
Case2-4

cr / yCase2-2
Case2-5

Case2-3

1) The flexural elastic buckling load for steel piles
in liquefied soil can be calculated from Eqs. (4)–(6) developed using the energy method.

1
0.82) Results clarified that piles’ dynamic buckling
occurs after soil liquefaction because of thesuperstructure weight and the superstructure
overturning moment.

0.6
0.4
× Case10.2
00○Results of static pushover analysis [3]Japanese design standard for steel structures[4]Japanese limit state design of steel structures[5] 

0.5

1

1.5

3) Pile dynamic buckling occurs when the bendingstrain reaches a threshold value.
4) The axial force after dynamic buckling for Case
2 series with soil decreases less than that of Case
1 series without soil.
5) The dynamic buckling stress of steel piles inliquefied soil can be estimated using themodified equivalent slenderness ratio bybuckling stress curves of the Japanese limit state
design of steel structures.

cFig. 29 Dynamic Buckling Strength and BucklingStress Curves.
dynamic buckling strength of piles on centrifuge
tests and the buckling stress curves for design
criteria in [4] and [5]. In Fig. 29, the curves
represent the buckling stress curves of the Japaneselimit state design of steel structures and the Japanesedesign standard for steel structures. The crosssymbols show results of Case 1. The other symbols
show results of Case 2. The small gray circle shows
results of static pushover analysis described in [3].
The results of centrifuge tests, except for Case 25, are exceeded for the curves of Japanese limit state
design of steel structures, distributed as the upperbound for the curve of Japanese design standard forsteel structures. For Case 2-5, which shows nodynamic buckling behavior, the maximum
compression axial force is substituted for the
dynamic buckling strength for reference.
Results show that dynamic buckling stress forthe steel piles in liquefied soil can be estimatedapproximately using the modified equivalent
slenderness ratio by the curve of the Japanese limit
state design of steel structures.

ACKNOWLEDGMENTS

The authors appreciate the support of the Japan
Iron and Steel Foundation.
REFERENCES[1] Architectural Institute of Japan, Preliminary
Reconnaissance Report of the 2011 TohokuChiho Taiheiyo-Oki Earthquake, July, 2011. (inJapanese)[2] Architectural
InstituteofJapan,
Recommendation for Design of Building
Foundations, 2001. (in Japanese)[3] Kimura Y, Tokimatsu K, “Buckling Stress of
Steel Pile with Vertical Load in Liquefied Soil”,
Jour.
of
StructuralandConstruction
Engineering, No. 595, Sep. 2005, pp.73-79. (inJapanese)[4] Architectural Institute of Japan, DesignStandard for Steel Structures, 2005. (inJapanese)[5] Architectural Institute of Japan, Limit State
Design of Steel Structures, 1998. (in Japanese)

CONCLUSION

Superstructure–pile foundation systems wereassessed using centrifuge tests to elucidate dynamic
buckling behaviors and buckling loads of circulartube piles under compressive axial forces. Results
show the following.

19

GEOMATE- Bangkok, Nov. 14-16, 2016

Technical Paper

20

GEOMATE- Bangkok, Nov. 14-16, 2016

Geotechnique21

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STABILITY CHARTS FOR UNSUPPORTED PLAIN STRAIN TUNNELHEADINGS IN UNDRAINED CLAY
Jim Shiau1, Brian Lamb1, Mathew Sams1, Jay Lobwein1 and Tin Nguyen1
School of Civil Engineering and Surveying, University of Southern Queensland, Australia

1

ABSTRACTThis paper investigates the stability of plane strain tunnel headings in undrained soils. Using the factor ofsafety approach via the strength reduction technique, the idealised tunnel heading models are studied using thefinite difference program FLAC. This problem is also similar to underground long wall excavations in planestrain condition. The finite difference results are presented alongside upper and lower bound limit solutions forvalidation. A thorough comparison between these two methods finds a very good agreement. Design charts arepresented for a wide range of practical scenarios using dimensionless ratios, similar to Taylor’s design charts
used for slope stability. Typical examples are presented to illustrate the usefulness for practicing engineers.
Keywords: Plain Strain Heading, Undrained Clay, Stability, Strength Reduction Method, Design Chart
INTRODUCTION
The research of tunnel heading stability was
initiated with [4] who studied the plastic flow of clay
soil in vertical openings such as sheet pile walls anddrew comparison to the stability of a tunnel heading
face. The stability number was initially defined in
Eq. (1).
𝑵𝑵 =𝝈𝝈𝒔𝒔 −𝝈𝝈𝒕𝒕 +𝜸𝜸(𝑪𝑪+𝑫𝑫/𝟐𝟐)
𝑺𝑺𝒖𝒖

All of these studies investigated the stability
number as a function of dimensionless parameters.
The problem was regarded as to find the limiting
value of a pressure ratio (σ s – σ t )/S u that is afunction of the independent parameters such as thedepth ratio C/D and the strength ratio S u /γD. It is
possible to simulate an unsupported excavation ingreen-field conditions by neglecting σ s and σ t , or asituation with equal surcharge and internal
supportive pressure. It can also be a factor of safety
problem that is a function of the depth ratio C/D andthe strength ratio S u /γD ~ a similar approach in
Taylor’s design charts for slope stability analysis[15].
This paper presents a comprehensive parametricundrained analysis for the stability of an
unsupported plain strain heading in an unsupported
excavation in green-field conditions. A strength
reduction technique is used to determine the factorof safety in cohesive soils over a wide parametricrange. Numerical results from the strength reduction
technique using FLAC are compared to thoseobtained by the finite element limit analysis. Design
stability charts are also presented for practical uses.

(1)

Where σ s is the uniform surcharge pressure on
the surface and σ t is the uniform internal tunnelpressure. The cover depth above the crown of thetunnel is C, whilst the tunnel diameter is D. The
undrained shear strength of soil and the unit weightof the soil are represented by S u and γ respectively.
During the 1970s at Cambridge University,
numerous studies were completed on centrifuge
models by [1]. The work was culminated by [9] who
investigated the experimental and theoreticalundrained collapse of three-dimensional cylindricaltunnel headings in normal consolidated kaolin under
different geometry and gravity regimes. References[5], [12] built further on the stability ratio by using
upper and lower bound solutions using the classicallimit theorems for tunnel collapse in undrainedcohesive conditions.
Reference [2] revisited the plane strain heading
problem due to the recent improvements to finite
element limit analysis methods developed in [7]-[8].
A number of papers were achieved in the area oftunnel stability by [10], [16] and [17]. In particular[10] extended the limit analysis theory on tunnel
heading by investigating a three-dimensional
multiblock failure mechanism for frictional andcohesive soil by the use of a spatial discretization
technique.

STATEMENT OF PROBLEM
The problem is a two dimensional plane strain
heading condition by assuming the longitudinal
section as a very long excavation [2]. Figure 1
shows the problem definition. The soil medium isconsidered as undrained and is modelled as a
uniform Tresca material, which is the same as a
Mohr-Coulomb material with no internal friction
angle. The undrained shear strength (S u ) and the unit
weight ( 𝛾𝛾 ) are soil properties used, whilst the22

GEOMATE- Bangkok, Nov. 14-16, 2016

excavation depth (D) has a cover depth (C) above
the crown of the excavation.

Fig. 2 Typical Mesh and Boundary Condition
The factor of safety for the many cases being
studied is computed by using FLAC and the built-in
implementation of the strength reduction technique.
A built-in FISH language that enables automaticmesh generation was also developed to improve theefficiency in the parametric study. FLAC is based on
the explicit finite difference method, but it is notvery different from a nonlinear finite element
program. This method has been widely used forslope stability analysis, but is rarely used for tunnel
stability analysis. This strength reduction techniquewill yield a factor of safety (FoS), which is not
unfamiliar with practicing designers.
Such a method involving factors of safety has
been described by [3] for slope stability. The factorof safety is defined as a ratio of the strength
necessary to maintain limiting equilibrium with thesoil’s available strength. The shear strength of thematerial is reduced until the limiting condition isfound. This can be compared to the studies
originated from [4] where the uniform surchargepressure is increased until the limiting condition isfound. If failure occurs initially, the shear strength of
the soil is increased by amplifying the cohesion andfriction angle until limiting equilibrium or failurestate is reached. Once the actual and critical strength
are known, the factor of safety can then be
calculated.

Fig. 1 Problem Definition
Similar to a slope stability analysis, the stabilityof the heading face is represented by the factor ofsafety (FoS) that is a function of the depth ratio(C/D) and strength ratio (S u /γD).
𝑪𝑪 𝑺𝑺

𝑭𝑭𝑭𝑭𝑭𝑭 = 𝒇𝒇 � , 𝒖𝒖 �
𝑫𝑫 𝜸𝜸𝜸𝜸(2)

These dimensionless ratios allow the results ofthis study to be used in scenarios that are physicallydifferent, but where the soil strength ratio and thedepth ratio still fall in the parametric domain. To
cover all possible realistic ranges, the parameters
used in the study include S u /γD = 0.1-2 and C/D = 16. This would ensure that the design charts produced
can be applied to many different tunnel design andanalysis problems, which are useful for design
purposes.

RESULTS AND DISCUSSION
Using the strength reduction technique within thefinite difference program FLAC, factor of safety(FoS) values were obtained for a range ofparameters for an unsupported tunnel heading inundrained soil. The two dimensionless parameters
used in the studies are the depth ratio (C/D) and thestrength ratio (S u /𝛾𝛾D).
It is important to compare the finite differencetechnique with another numerical investigation.
Using finite element limit analysis, upper bound andlower bound, factors of safety have been calculatedusing OptumG2 over the same parametric ranges.
The numerical procedures used in OptumG2 arebased on the limit theorems of classical plasticity[7]-[8].

MODELLING PROCEDURE
Figure 2 shows a typical finite difference meshof the problem in this study. The soil domain size
was elected to be large enough to validate theassumption of an infinite excavation. The boundary
conditions within Fig. 2 are important to ensure thatthe entire soil mass is modelled accurately despiteusing a finite mesh. The smooth rigid lining aboveand below the soil excavation is restrained in thevertical (y) direction to reproduce the nature of thetunnel linings and mining supports. The base andsides of the model is restrained in the x and y
directions.

COMPARISONS WITH RIGOROUS UPPER
AND LOWER BOUNDS
Dimensionless stability charts showing thenumerical results obtained in this study are presented
in Fig. 3 and 4. In Fig. 3, FoS increases linearly as
the strength ratio S u /γD increases, indicating that
there exists a stability number where the effective
FoS is equal to one, a critical stability number (N c ).
This could be achieved by dividing the stability ratio
23

GEOMATE- Bangkok, Nov. 14-16, 2016

(S u /γD) by the FoS result for each case. Also notethat the rate of FoS increase is different for each C/Dvalue. The gradient of the line is greater for smaller
C/D values.

C/D=1

cohesive soil but at different depths, the tunnel withthe smaller active force (γC) will have a higher
probabililty of stability. This observation may not
be true in a soil with internal friction angle (ϕ≠ 𝟎𝟎)
due to the additional shear strength (σtanϕ) andgeometrical arching effects. In purely cohesive soils,
the latter still occurs, but its effect is not enough toovercome that subsequent increase in active force.

C/D=2

EXTENT OF FAILURE SURFACE

7
Lower Bound, Optum G2, This paper
Upper Bound, Optum G2, This paper
Shear Strength Reduction, FLAC, This paper

6

5

FoS

4
C/D=3

This area of investigation is related to the type of
failure mechanism the tunnel heading encountered atcertain depths. Reference [2] discusses the effects ofincreasing the soil weight and linear variation of theshear strength of the soil for increasing depth ratios.
For homogenous shallow tunnels (depth ratios C/D ≤2), the failure mode is predominantly through thevertical free movement wall. A trapdoor typemechanism is initiated, with the velocity vectors
leading vertically down above the heading opening.
For deeper tunnels the failure zone extends to thebase of the tunnel. The velocity vectors are archedaround the free moving heading vertical surface,compared to the vertical trapdoor mechanism for theshallow tunnels. Figures 5-8 show some typical plotsof the shear strain rate and velocity field. Note thatthe cohesive strength of the soil appears to have no
impact on the failure extent.

C/D=4
C/D=5
C/D=6

3

2

1

0
0.0

0.5

1.0

1.5

2.0

2.5

S u/ D

Fig. 3 Comparison of FoS results with respect toS u /γD for various values of C/D
6
Lower Bound, Optum G2, This paper
Upper Bound, Optum G2, This paper
Shear Strength Reduction, FLAC, This paper

5

FoS

4

3

Su/ D = 2.0

21

Su/ D = 1.3
Su/ D = 0.9

0Su/ D = 0.5
Su/ D = 0.1
0

1

2

3

4

5

6

7

C/D

Fig. 4 Comparison of FoS results with respect toC/D for various values of S u /γDFigure 4 shows that the FoS decreases
nonlinearly with increasing depth ratio C/D for all
strength ratios defined as S u / 𝛾𝛾 D. The finite
difference results are in good agreement with theupper bound solutions, but consistently produce
results that are 3-5% larger than the lower boundsolutions. Note the strength ratio is normalised withrespect to the tunnel excavation depth (D), andtherefore the undrained shear strength (S u ) remains
constant throughout the increasing depth ratios.
Owing to the increasing overburden pressure (C/D
increase) and the constant undrained shear strength(S u ), it therefore results in FoS values decreasing.
This is in contrast to the common belief that anincrease to C/D always results in an increase to FoS.
A simple observation can be made from Fig. 1,
where the active force is the weight of soil and theresisting force is given by the shear strength of thesoil. Of two hypothetical tunnels in the same

Fig. 5 Plots of shear strain rate and velocity field for
C/D=1 and S u /γD =0.1

Fig. 6 Plots of shear strain rate and velocity field for
C/D=1 and S u /γD =2.024

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 9 Principle stress tensor plot at collapse for
C/D=1 and S u /γD =2.0

Fig. 7 Plots of shear strain rate and velocity field for
C/D=3 and S u /γD =0.1

Fig.10 Principle stress tensor plot at collapse for
C/D=3 and S u /γD =2.0
In reality the deformation trough would be three
dimensional, but with the assumptions made, thelargest extent of plastic deformation in the heading
cross-section can be reasonably estimated. This
information of failure extent is important as it willassist practicing engineers in the decision making inmonitoring ground movements.
USING THE STABILITY CHARTS
The stability design charts are best demonstrated
through a number of examples. The numericalresults have been used to produce design contour
charts for factors of safety in Fig. 11. A designer
will use the chart to relate the depth ratio (C/D), soil
strength ratio (S u /𝛾𝛾D) and factor of safety for theparticular tunnel heading. This process would
predominantly be used in the initial design
procedure for the design engineers.
Fig. 11 presents the design chart where the factorof safety is a function of the depth ratio and strength
ratio. Regression of the chart gives the following
relationship with r2 = 0.99.

Fig. 8 Plots of shear strain rate and velocity field for
C/D=3 and S u /γD =2.0
Figures 9 and 10 show typical tensor plots forprincipal stresses. These plots show weak soil
arching throughout the soil body and leading to thesurface from the free excavation face. As discussed,
soils with an internal friction angle (ϕ≠ 0) would
have more potential for stability, with the internal
frictional angle adding to the strength of the material
by the soil arch phenomena.

𝒔𝒔

𝑪𝑪 −𝟎𝟎.𝟒𝟒𝟒𝟒

𝑭𝑭𝑭𝑭𝑭𝑭 = 𝟑𝟑 � 𝒖𝒖 � � �
𝜸𝜸𝜸𝜸

25

𝑫𝑫

(3)

GEOMATE- Bangkok, Nov. 14-16, 2016

An actual computer analysis of this particularcase gives a FoS of 0.51. Therefore, in this case, afailure to supply heading pressure would result incollapse. Ground improvement is needed to increase
the soil strength.

2.0
5.5
4.0

5.0
1.8

2.5

3.0
3.5

4.5

1.6

2.5

3.0

2.0

4.0
3.5

1.4

S u/ D

2.5
1.2

3.0

1.0

2.5

0.8

2.0

Design of an unsupported tunnel heading
The soil properties are known at the tunnel
project site, and the diameter is specified. A target
factor of safety is chosen, and the designers need to
specify a maximum cover depth that will satisfy thetarget FoS.
Parameters are given as: S u = 100 kPa, 𝜸𝜸 = 18 kN/m3,
D = 5m, and the target FoS = 2

2.0
1.5

2.0
1.5
1.5

1.0
1.0

0.6

1.5

0.4

1.0

1.0
0.5
0.5

1.

0.5
0.2

0.5
1

2

3

4

5

2.

6

C/D

Fig.11 Stability chart for FoS with respect to C/D
and S u /γD

Using FoS = 2 and S u /γD = 1.11, Eq. 3 gives a
C value of 14.8 m.
Using FoS = 2 and S u /γD = 1.11, Fig. 11 gives
an approximate C/D value of 3 and therefore C
value of 15 m.

An actual computer analysis for this particular case(C value of 15m) gives a FoS of 2.07

EXAMPLESCONCLUSION

Using the design chart and the regressed design
equation, practical examples are presented for either
analysis or design purposes.

Stability of plane strain headings was studied in thispaper using a factor of safety approach with thestrength reduction technique. Numerical results wereobtained using both the finite difference software
FLAC and the finite element software OptumG2.
Design equation and charts were produced alignedwith examples to determine the stability of
excavated heading depths, such as idealised tunnel
headings and long wall mining excavations. The
factor of safety approach to plain strain heading
stability is a useful approach to the initial design
stage as it always provides direct information andunderstanding of tunnel stability.

Analysis of an existing unsupported tunnel
An existing mining excavation having no internal
heading pressure and no surcharge pressure,determine the factor of safety for operations tocontinue. Parameters are given as S u = 50 kPa, γ =18 kN/m3, C = 10 m, and D = 2.5 m.
1.
2.

Using C/D = 4.0, S u /γD = 1.11, Eq. 3 gives a
FoS of 1.74.
Using C/D = 4.0, S u /γD = 1.11, Fig. 11 gives anapproximate FoS of 1.75.

REFERENCES

An actual computer analysis of this particular casegives a FoS of 1.76.

[1] Atkinson, JH, & Cairncross, AM 1973,‘Collapse of a shallow tunnel in a Mohr–
Coulomb material’, In Proceedings of the
Symposium on the Role of Plasticity in Soil
Mechanics, Cambridge, UK, 13–15 September
1973. Edited by A.C. Palmer. Cambridge
University Engineering Department, Universityof Cambridge, Cambridge, UK. pp. 202–206.
[2] Augarde, CE, Lyamin, AV & Sloan, SW 2003,‘Stability of an undrained plane strain heading
revisited’, Computers & Geotechnics, vol.30, pp.
419-430.
[3] Bishop, AW 1955, ‘The use of slip circle in thestability analysis of slopes’, Geotechnique, vol.
5(1), pp. 7-17.
[4] Broms, BB & Bennermark, H 1967, ‘Stabilityof clay at vertical openings’, Journal of the Soil

Analysis of a temporary unsupported tunnel heading
During the construction of a tunnel in soft soil, afactor of safety of 3 is targeted for a shallow tunnel
maintenance job. While the machine requires
maintenance, it will be unable to apply pressure to
the tunnel face for a short period. A decision needs
to be made whether stability will be maintained
during this period. Parameters are given as: S u = 30
kPa, 𝜸𝜸 = 18 kN/m3, C = 20 m and D = 5 m
1.

2.

Using C/D = 4.0, S u /γD = 0.33, Eq. 3 gives a
FoS of 0.52.
Using C/D = 4.0, S u /γD = 0.33, Fig. 11 gives anapproximate FoS of 0.54.

26

GEOMATE- Bangkok, Nov. 14-16, 2016

Mechanic & Foundations Division, Proceedingsof the American Society of Civil Engineers, vol.
93, pp. 71- 93.
[5] Davis, EH, Gunn, MJ, Mair, RJ & Seneviratne,
HN 1980, ‘The stability of shallow tunnels &underground openings in cohesive material’,
Geotechnique, vol. 30, pp. 397-416.
[6] FLAC 2003, Fast Lagrangian Analysis ofContinua, Version 4.0, Itasca Consulting Group,
Minneapolis, Minnesota, USA.
[7] Lyamin, AV, & Sloan, SW 2002(a), ‘Lowerbound limit analysis using non‐linear
programming’, International
Journalfor
Numerical Methods in Engineering, vol. 55, no.
5, pp. 573-611.
[8] Lyamin, AV & Sloan, SW 2002(b), ‘Upper
bound limit analysis using linear finite elements&
non‐linear
programming’, International
Journal for Numerical & Analytical Methods inGeomechanics, vol. 26, no. 2, pp. 181-216.
[9] Mair RJ 1979, ‘Centrifugal modelling of tunnel
construction in soft clay’, PhD thesis,
University of Cambridge.
[10] Mollon, G, Dias, D & Soubra, AH 2010, ‘Face
Stability Analysis of Circular Tunnels Driven
by a Pressurized Shield’, Journal of
Geotechnical & Geoenvironmental Engineering,
Vol: 136, pp. 215-229.
[11] Optum G2 2013, Optum Computational
Engineering, Version 1.2014.10.1, Newcastle,
Australia.
[12] Sloan, SW, Assadi, A 1994, ‘Undrained
stability of a plane strain heading’, Canadian
Geotehcnical Journal, vol. 31, no. 3, pp. 443450.

[13] Sloan, SW 1988, ‘Lower bound limit analysis
using finite elements & linear programming’,
International Journal for Numerical &Analytical Methods in Geomechanics, vol. 12,
pp. 61-67.
[14] Sloan, SW 1989, ‘Upper bound limit analysis
using finite elements & linear programming’,
International Journal for Numerical &Analytical Methods in Geomechanics, vol. 13,
pp. 263-282.
[15] Taylor, DW 1937, ‘Stability of earth slopes.’
Journal of the Boston Society of Civil Engineers,
vol. 24, no. 3, pp. 197-246.
[16] Wilson, DW, Abbo, AJ, Sloan, SW, & Lyamin,
AV 2014, ‘Undrained Stability of Dual Circular
Tunnels’,
International
Journalof
Geomechanics, vol. 14(1), pp. 69-79.
[17] Yamamoto, K, Lyamin, AV, Wilson, DW,
Sloan, SW & Abbo, AJ 2011, ‘Stability of acircular tunnel in cohesive-frictional soil
subjected to surcharge loading’, Computers &Geotechnics, vol. 38, pp. 504-514.
[18] Shiau, J., Sams, M. and Chen, J. 2016, ‘Stability
charts for a tall tunnel in undrained clay’, Int. J.
of GEOMATE, vol. 10(2), pp. 1764-1769[19] Shiau, J., Sams, M. and Lamb, B. 2016,‘Introducing advanced topics in geotechnicalengineering teaching – Tunnel modelling’, Int. J.
of GEOMATE, vol. 10(1), pp. 1698-1705[20] Shiau, J., Lamb, B. and Sams, M. 2016, ‘The
use of sinkhole models in advanced
geotechnical engineering teaching’, Int. J. of
GEOMATE, vol. 10(2), pp. 1718-1724

27

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STRENGTH AND PERMEABILITY CHARACTERISTICS OF ROAD
BASE MATERIALS BLENDED WITH FLY ASH AND BOTTOM ASH
1,2

Jonathan R. Dungca1 and Julie Ann L. Jao2
Civil Engineering Department, De La Salle University, Manila, Philippines

ABSTRACTThe Philippines has an extensive road network which handles most of its passenger and freight movements.
Large volumes of aggregate embankment materials of good quality are required to primarily support thesetransport infrastructures, and this poses threat to the environment. Coal combustion by-products (CCPs) are seen
to be its potential alternative mainly due to its vast production and disposal problems in the country.
Representative samples of class C fly ash and bottom ash were gathered together with conventional road basematerials. Fly ashes were substituted to act as fines; whereas, bottom ash substitutions were varied at different
mixture ratios of 0%, 20%, 40%, 60%, 80%, and 100% of fine aggregates. Index properties (i.e. specific gravity,
Atterberg limits, and maximum and minimum index densities), compaction characteristics, unsoaked and soaked
California Bearing Ratios (CBR), and hydraulic conductivities were obtained for all the blends in order toproduce empirical relationships with varying bottom ash content. Results show that the optimum strength can beproduced at a blend of 100% bottom ash. However, permeability tests show a considerable decline in hydraulicconductivity with the addition of coal ashes to the typical aggregates. Thus, proper drainage must be carefully
applied to these blended embankment materials so as to avoid substantial ingress of water.
Keywords: Fly Ash, Bottom Ash, Road Embankment, CBR, Permeability

INTRODUCTION

The Particle Size Analysis was performed inaccordance with ASTM D422 [3]. It determines thequantitative distribution of particle sizes of fly ash,bottom ash, and conventional materials used in the
study. Sieve analyses were done for particle sizes
larger than 75 μm (or those retained at No. 200
sieve); while hydrometer analyses were conducted
for particle sizes smaller than 75 μm (or thosepassing No. 200 sieve).
Conventional materials, fly ash, and bottom ash
samples were individually subjected to microscopic
examinations in order to characterize their particle
angularity, assemblage, and surface texture. These
were performed using Scanning Electron
Microscopy (SEM). With the purpose ofcharacterizing the chemicals present on the
introduced materials and their corresponding
proportions, Energy-Dispersive X-ray Spectroscopy(EDS) was performed to both coal ashes. All threematerials were subjected to X-ray Diffraction (XRD)
Analysis so to investigate their structures.
To provide initial results, the following Index
tests were conducted on all the pure materials
conforming to the ASTM procedures.
a. Specific Gravity of Soils (ASTM D854)
b. Atterberg Limits (ASTM D4318 and D427)
c. Maximum and Minimum Index Densities(ASTM D4253 and D4254)
These tests were also performed on the blended
materials while keeping its grain size distribution
constant. It should be noted, however, that the grain

Transport infrastructure plays an important rolein integrating the island economies of an archipelago
such as the Philippines. Mostly comprising the
highway embankment materials are the so-calledaggregates. Due to the increasing public attention
being paid to the environment, the coal combustion
by-products (CCPs) are often looked upon as an
attractive alternative to these aggregates mainly due
to its large production and disposal problems. In fact,
the coal-fired power plants which produce
considerable amount of CCPs contribute the largest
share of 27.06% of the total installed capacity of
15,571 MW last 2009 [1].
The main objective of this paper is to determine
the optimum blending proportion of fly ash andbottom ash to the conventional road base materials
used as highway embankments. Aside fromillustrating the morphological characteristics of thematerials in their pure forms, emphasis is given to
the determination of their geotechnical properties
when blended at different mixture ratios.
METHODOLOGYAsh Properties
Using ASTM D854, the specific gravity of eachsoil blend was determined. The specific gravity ofthe soil mixtures was reduced by the addition of fly
ash [2] since the usual of the specific gravity of fly
ash is much lower compared with the soil.
28

GEOMATE- Bangkok, Nov. 14-16, 2016

size distribution curve (GSDC) used for all the
specimens were arithmetically computed to comply
with the requirements stipulated by Department of
Public Works and Highways (DPWH) for sub-baseand base courses. With this, fly ash was used to
substitute the entire fines content constituting 10%of the total mass. On the other hand, the fineaggregates which comprise 32.5% of the total mass
were replaced by bottom ash at different mixture
ratios of 0%, 20%, 40%, 60%, 80%, and 100%.

100

ActualGradation Used80
% Finer

Lower Limit of
DPWH
Requirement

60

Upper Limit of
DPWH
Requirement

40
20
0
100

Mechanical Properties
Following ASTM D698, Standard Proctor Test
determined the OMC at which dry unit weight isgreatest and compaction is best. The OMC that wereobtained from these tests were used for sample
preparation on strength and permeability tests foreach blend.
Water was added to the sample to prepare thespecimens such that the moisture contents are closed
to 100% of OMC obtained in the Standard Proctor
Test. After compaction, the compacted specimen in
the mould was trimmed level with the top surface
and then inverted to have the previously bottom
surface tested under the Uniframe. CBR tests(ASTM D1883) were conducted under soaked andunsoaked conditions. The penetration test was
performed immediately after the compaction forunsoaked condition.
Falling Head Permeability Tests (ASTM D5084)
were conducted so as to illustrate the drainage
behavior of all the blends under Relative
Compaction, RC=100%. However, since the desired
RC=100% is somehow unattainable in the laboratory
due to tamping constraints, tests were instead
conducted under varying relative compactions (i.e.
80%, 85%, and 90%) to produce a reliable empiricalmodel.

1
Grain Diameter
(mm)

0.01

Fig. 1. Actual Grain Size Distribution Curve Used
for All Tests
Table 1. USCS and AASHTO Soil Classifications ofPure Materials
AASHTO
USCS(Group Index
Material(ASTM D2487)
Method)Well-Graded
Conventional
Gravel with Silt
Gravel andMaterials
and Sand (GWSandGM)
Fly AshSilt (ML)
Silty Soils
Poorly Graded
Bottom Ash
Sand with Gravel
Fine Sand(SP)
Microscopic Examination
Apparently, its structure is composed of coarse
and fine particle sizes with a few silt-sized ones
shown on Figure 2. This supports the result of its
soil classification which was “well-graded gravelwith silt and sand”. However, at higher
magnification, flakes at random direction weredetected to comprise its grains; while few interassemblage pore spaces were also noted.

RESULTS AND DISCUSSIONS
Gradation
It can be seen on Figure 1 that 57.5% of thetotal mass is composed of coarse aggregates,
whereas 32.5% are fine aggregates and 10% arefines.
Fly ash is predominantly composed of silt- toclay-sized particles or those finer than 0.075 mm(No. 200 sieve). On the other hand, bottom ash has abetter grain size distribution than the former withmostly sand-sized particle composition.

(a)(b)
Fig. 2. SEM Photomicrographs of Conventional
Road Base Materials at (a) 100X and (b) 5000X
Magnifications

Soil Classification
Shown on Table 1 are the results of soil
classificationaccordingtoUnified
Soil
Classification System (USCS) [4, 5]. These werethen verified using the AASHTO Soil Classification
System following the Group Index Method.

Fly ashes are composed generally of sphericalsilt-sized particles with smooth surface as shown on
Fig. 3. It can be seen that they flocculate with one
another thereby producing larger and denser
clustered particles. This tends to result in highvolume of voids. Hence, at higher magnification,
large inter-granular voids can be observed. This
29

GEOMATE- Bangkok, Nov. 14-16, 2016validates the high air void content, associated with
its compaction behavior.

Table 2. EDS Results of Fly Ash and Bottom Ash
Elements
Fly AshBottom Ash
Al
14.58
0.00
Si
30.37
6.83
K
4.77
0.00
Ca
6.58
16.40
Fe
2.71
2.66
Rb
34.29
7.54
Mo
1.06
0.00
I
1.10
2.28
Ba
1.44
0.88
Np
3.10
0.00
O
0.00
44.04
Mg
0.00
1.49
Sb
0.00
10.37
Er
0.00
7.51

(a)
(b)
Fig. 3. SEM Photomicrographs of Fly Ash at (a)2000X and (b) 5000X Magnifications
As presented in Fig. 4, comparing its shape andsurface characteristics, bottom ashes are seen
comparable to the conventional materials in terms ofparticle sizes. However, the large grains can beobserved to have clothed contacts since some of thesilt-sized particles attach themselves to the muchcoarser grains. Few inter-assemblage pore spaces
can also be observed with the 5000X magnification.

Although similar in size and behavior to naturalsand, the chemical composition of bottom ash
provides unique pozzolanic properties that, as withcementitious materials, can result in a favorable
time-dependent increase in strength. This is mainly
due to its high calcium content which tends to reactwith water from which pozzolanic reaction occurs.
Geotechnical Properties of Pure Materials
Relevant geotechnical tests conforming to
ASTM were individually conducted on all threematerials in order to come up with the results, as
shown on Table 3.
The obtained specific gravity of the aggregate
samples is about 2.81, which is within the typicalrange of 2.40 to 3.60 depending on the nature of themineral constituents. It can be seen that coal ashes
have much lower specific gravities than theconventional aggregate materials. Bottom ashes
have higher specific gravity than fly ash owing to
the presence of heavier particles. It appears thatashes having high iron oxide content may have highspecific gravity [6].
Atterberg Limits Test using the Percussion Cup
Method was used to obtain the liquid limit (LL) of
15.60 and plasticity index (PI) of 1.96 forconventional road base materials. Both are within
the restrictions of sub-base and base courses as per
DPWH requirements.
With the results from the three materials, it isevident that coal ashes are indeed lighter than theconventional materials and is supported by the
specific gravities obtained prior. The void ratio of asoil is largely associated with particle size
distribution, particle size, shape, and texture. Lower
maximum and minimum void ratios were observed
for conventional materials; thus, indicating that it
has better distribution of particle sizes.
Results demonstrate that fly ash has lessvariation of dry density compared to that for a wellgraded soil. This is why fly ash is allowed to be(a)(b)
Fig. 4. SEM Photomicrographs of Bottom Ash at (a)100X and (b) 5000X Magnifications
Chemical Analysis
The chemical properties of the coal ashes
greatly influence the environmental impacts that
may arise out of their use/disposal as well as theirengineering properties. The adverse impacts include
contamination of surface and subsurface water with
toxic heavy metals present in the coal ashes, loss ofsoil fertility around the plant sites, and the like. In
practice, EDS is most often used for qualitative
elemental analysis, simply to determine whichelements are present on the materials and theirrelative abundance.
The bulk composition of fly ash is similar to
many geologic materials. Fly ash is primarily
composed of silicon, aluminum, iron, calcium,
potassium, and rubidium, associated with oxygen asoxides, silicates, and aluminates. The combined
silicon, aluminum, and iron content (reported asoxides) are frequently used to provide an indicationof the pozzolanic or cementitious nature of fly ash.
A combined value of 70% of these components
indicates a pozzolanic fly ash, while a value of
between 40% and 70% indicates a cementitious fly
ash. Based on the results of EDS, fly ash samples areto be considered cementitious because the total
content of oxides is 47.66%.
30

GEOMATE- Bangkok, Nov. 14-16, 2016

Specific Gravity, Gs

compacted over a larger range of water content [7].
Furthermore, because of the generally low specificgravities of coal ash compared to soils, ash fills tend
to result in lower dry densities. This decrease in the
MDD is due to the extensive agglomeration of ash
particles that prohibits the specimens to becompacted properly. Soils can be stabilized
effectively by cation exchange using fly ash [8].
There are laboratory experiments using local
samples [9, 10, 11, 12] of bottom ashes that can beused for validation. A noticeable decrease in CBRstrength was manifested when the specimens were
tested after soaking it in water for four (4) days [13].

2.90

Specific Gravity

2.85

Upper Limit

2.80

Lower Limit

2.75 2.780

Average Specific Gravity

2.745

2.70
2.65

2.695

2.60
2.572

2.55
R² = 0.951

2.50

2.519

2.45
2.40

Table 3. Geotechnical Properties of Pure Materials
Conventional
Fly
BottomProperty
Materials
Ash
Ash
2.813
2.335
2.519
Specific Gravity, G s
15.60
Liquid Limit, LL
1.96
Plasticity Index, PI
Minimum Index
16.9605.614
11.785
Density,3
γ dmin (kN/m )Maximum Index
22.2898.330
16.869Density,γ dmax (kN/m3)
Optimum Moisture
4.938
46.16414.058Content, OMC (%)
Maximum Dry
23.621
10.18417.298Density, MDD(kN/m3)43.05%
48.83%
30.27%
Unsoaked CBR
33.60%
Soaked CBRHydraulic
Conductivity, k
4.58E-4
4.35E-5
1.04E-5(cm/sec) at
RC=100%
Falling Head Permeability Tests wereconducted at different relative compactions of
RC=80%, 85%, and 90% in order to illustrate theeffect of varying void ratios. Smaller particles comewith smaller voids between them; and hence
concluded that the resistance to flow of water
increases with decreasing size of particles. As aresult, the permeability decreases. Also, particles
with a rough surface texture provide more frictional
resistance to flow than smooth-textured particles.

20

40

60

80

100

Bottom Ash Content, B (%)

Fig. 5. Specific Gravity vs. Bottom Ash Content
Maximum and Minimum Index Densities ofBlended Materials
To see and understand its behavior with varying
bottom ash content, Equations 1 and 2 was generated.
As observed, there is a significant decrease in both
γ dmin and γ dmax when fly ash is present in mixture
with aggregates, this is due to the significantly
lighter weight of fly ashes which are hollow
particles that were used to substitute the fines
content of natural soil to complement the study ofKim [13].
Further addition of bottom ash tends to increase
the values of both γ dmin and γ dmax . The addition ofbottom ash leads to increasingly more well-gradedsize distributions, which allows the fly and bottom
ash particles to pack more closely with the coarse
aggregates, resulting in the increase in γ dmax .
(1)
𝜸𝜸𝒅𝒅𝒅𝒅𝒅𝒅𝒅𝒅 = 𝟏𝟏𝟏𝟏. 𝟔𝟔𝟔𝟔𝟔𝟔𝟔𝟔𝒆𝒆𝒆𝒆𝒆𝒆𝟎𝟎.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎(2)𝜸𝜸𝒅𝒅𝒅𝒅𝒅𝒅𝒅𝒅 = 𝟐𝟐𝟐𝟐. 𝟒𝟒𝟒𝟒𝟒𝟒𝟒𝟒𝒆𝒆𝒆𝒆𝒆𝒆𝟎𝟎.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎
where: γ dmin = minimum index density;
γ dmax = maximum index density; andB = bottom ash content (%)
Compaction Behavior of Blended Materials
In summary, Fig. 6 shows the behavior of the
OMC with increasing bottom ash content. As
observed, it shows a sudden increase in OMC when
fly ash was introduced to the mixture to act as its
fines content passing No. 200 sieve. This is mainly
due to the capability of fly ash to contain large
amount of water because of its high air void content
nature. Further increase in bottom ash content
slightly increases the OMC owing to the same
reason. Moreover, an abrupt decrease in MDD was
also noted with the use of fly ash because of its
much lighter weight and the agglomeration of its
particles prohibiting the specimens to be properly
compacted.

Geotechnical Properties of Blended Materials
Specific Gravity of Blended Materials
Since specific gravity (G s ) is one of the mostimportant physical properties required incharacterizing the usability of coal ashes forgeotechnical and other applications, it was obtained
for all the blending proportions as shown on Fig. 5.

31

GEOMATE- Bangkok, Nov. 14-16, 2016

OMC

120

Upper Limit

10

Lower Limit

9.5

Average OMC

Corrected Unsoaked CBR (%)

Optimum Moisture
Content,OMC (%)

10.5

9

8.5
8

7.5 7.979
7

8.138

8.279

8.402

8.567

6.5
6
20

40
60
80
Bottom Ash Content, B (%)

Unsoaked
CBR

110
R² = 0.710

Upper
Limit

100
90
80

88.36

88.90

86.04

70

AverageUnsoaked
CBR
Power(Unsoaked
CBR)

71.57

60
56.32
50

100

40
20

Fig. 6. Optimum Moisture Content (OMC) vs.
Bottom Ash Content

40

60

Lower
Limit

80

100

Bottom Ash Content, B (%)

Fig. 7. Unsoaked CBR Values vs. Bottom Ash
Content

The empirical equations correlating OMC and
MDD with the bottom ash content are respectively
illustrated in Equations 3 and 4.

250

𝟎𝟎.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎

Corrected Soaked CBR (%)(3)
𝑶𝑶𝑶𝑶𝑶𝑶 = 𝟕𝟕. 𝟖𝟖𝟖𝟖𝟖𝟖𝟖𝟖𝒆𝒆𝒆𝒆𝒆𝒆
𝑴𝑴𝑴𝑴𝑴𝑴 = 𝟐𝟐𝟐𝟐. 𝟕𝟕𝟕𝟕𝟕𝟕𝟕𝟕𝒆𝒆𝒆𝒆𝒆𝒆−𝟎𝟎.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎 (4)where: OMC = optimum moisture content (%);
MDD = maximum dry density (kN/m3);
and
B = bottom ash content (%)
Strength Behavior of Blended Materials
Presented on Figs. 7 and 8 are the corrected
CBR values obtained for unsoaked and soaked
conditions, respectively. These were plotted toclearly illustrate its behaviors with increasing
bottom ash content while fly ash percentage was
fixed at 10% of total mass.
It can be observed that there is a significant
increase in CBR strength when fly ash was used to
substitute the entire fines content constituting 10%of the total mass. The strength of fly ash generally
improves with time due to cementitious reactions.
Reactive silica and free lime contents are necessary
for this reaction to take place. Given all these data,
empirical relationships expressed in Equations 5 and6 were established with which unsoaked and soaked
CBR values can be respectively calculated at any
given bottom ash content.
As affirmed, a continuous increase in CBR withincreasing bottom ash content lead to a conclusionthat the blend F10-B32.5-C57.5, in which 100% offine aggregates are substituted by bottom ash, is the
optimum blend which provided the highest CBRvalues. This blend provided a notable unsoaked
CBR of 88.90% and a remarkable soaked CBR of
212.87%, indicating its strength to be more than
twice that of crushed rock.

SoakedCBR

R² = 0.957

200

Upper
Limit

212.87
192.31
150

Lower
Limit

158.05130.77100

Average
Soaked
CBR103.81
50
20

40

60

80

100

Bottom Ash Content, B (%)

Fig. 8. Soaked CBR Values vs. Bottom Ash Content
𝑪𝑪𝑪𝑪𝑪𝑪𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖𝒖 = 𝟐𝟐𝟐𝟐. 𝟖𝟖𝟖𝟖𝟖𝟖𝟖𝟖𝑩𝑩𝟎𝟎.𝟐𝟐𝟐𝟐𝟐𝟐𝟐𝟐 (5)𝑪𝑪𝑪𝑪𝑪𝑪𝒔𝒔𝒔𝒔𝒔𝒔𝒔𝒔𝒔𝒔𝒔𝒔 = 𝟐𝟐𝟐𝟐. 𝟕𝟕𝟕𝟕𝟕𝟕𝟕𝟕𝑩𝑩𝟎𝟎.𝟒𝟒𝟒𝟒𝟒𝟒𝟒𝟒 (6)where: CBR unsoaked = unsoaked CBR value (%);
= soaked CBR value (%); andCBR soaked
B = bottom ash content (%)
Permeability Behavior of Blended Materials
Falling Head Permeability Tests in this study
were conducted only under varying relative
compactions of RC=80%, 85%, and 90% because
samples with RC=95% and higher were found to beunattainable since specimens were only to be tamped
using hands. Shown on Fig. 9 are the raw dataobtained for all the blends.
𝒌𝒌 = (𝟐𝟐. 𝟗𝟗𝟗𝟗𝟗𝟗𝟗𝟗𝟗𝟗−𝟕𝟕 )𝒆𝒆𝒆𝒆𝒆𝒆(𝟏𝟏𝟏𝟏.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎 + 𝟎𝟎.𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎)

where: k = hydraulic conductivity (cm/sec);e = void ratio of the material; andB = bottom ash content (%)32(7)

GEOMATE- Bangkok, Nov. 14-16, 2016

Due to the large specific surface of fly ash, itcauses more resistance to flow of water through thevoids. With this, the addition of fly ash in the
mixture resulted to a sudden decrease inpermeability compared to the controlled mix which
has a hydraulic conductivity of 4.58x10-4 at
RC=100%.

Hydraulic Conductivity, k(cm/sec)

1.00E-02
20% B
40% B

1.00E-03

60% B
80% B

1.00E-04100%
1.00E-05
0.2

0.3

0.4

0.5

0.6

Void Ratio, eFig. 9. Hydraulic Conductivity vs. Void Ratio
Subsequently, the results were compared to thedrainage characterization as presented by Hazelton
and Murphy [15] wherein it was found that all of theblended materials were categorized to have “veryslow infiltration” at RC=100%.
CONCLUSIONS AND RECOMMENDATIONS
After performing all the necessary geotechnicaltests, it can be concluded that both class C fly ash
and bottom ash produced from burning local coals
can be safely utilized as road embankment materials
together with the natural aggregates typically used.
Also, the self-cementing and pozzolanic hardening
properties of class C fly ash and bottom ash impart
additional strength to the road embankments. Since
percolation of water through the soil is highly
anticipated especially during rainy season,
cementitious reaction can be expected to take place
thereby strengthening the blended materials. The
addition of bottom ash from 0% to approximately
50% of fine aggregates gave increasing hydraulicconductivity because they act just like the natural
sands wherein water can flow freely. Ultimately, theutilization of coal ashes in highway embankment not
only diminishes construction cost, but also helps indisposal problem of these previously regarded aswaste by-products.
REFERENCES[1] Department of Energy. (2010). Philippine
Energy Plan: 2009 Update. Retrieved February
10, 2012 from http://www.doe.gov.ph/PEP/
33[2] Prabakar, J., Dendorkar, N., & Morchhale, R.
(2004). Influence of fly ash on strength behaviorof typical soils. Construction and Building
Materials, 18, 263-267.
[3] American Society for Testing and Materials.
(2004). ASTM Standards on Transportation
Applications. American Soc. for Testing & Mat.
[4] U.S. Army Military Soils Engineering (1997).
Field
Manual5-410.
http://www.adtdl.army.mil/cgi-bin/atdl.dll/fm/5410/toc.htm[5] U.S. Army Military Soils Engineering (1997).
Field
Manual5-410.
http://www.itc.nl/~rossiter/Docs/FM5410/FM5-410_Ch12.pdf[6] Seals R.K., Moulton L.K., and Ruth, B.E.
(1972). Bottom Ash: An Engineering Material.
JSMFD, ASCE, 98, 311-325.
[7] Pandian, N.S. (2004). Fly Ash Characterization
with Reference to Geotechnical Applications.
Indian Institute of Science, 84, 189-216.
Retrievedfrom
http://journal.library.iisc.ernet.in/[8] Bhuvaneshwari, S., Robinson, R., & Gandhi, S.
(2010). Stabilization of Expansive Soils Using
Fly Ash. Fly Ash Bricks.
[9] Marques, J., Arvesu, M., & Aguila, B. (2003).
A Study on Bottom Ash as Substitute for Fine
Aggregates in the Production of Reinforced
Pipe Culverts. (Undergraduate Thesis) De La
Salle University, Manila, Philippines.
[10] Antonio, O. (2006). Lime-Fly Ash Stabilizationof Calaca, Batangas Bottom Ash. (Masteral
Thesis) University of the Philippines, Quezon
City, Philippines.
[11] Asilo, M., Pamintuan, C., & Roque, A. (2004).
Bottom Ash By-Product from Coal-Fired Power
Plant in Pangasinan Used as a Substitute forFine Aggregates in Concrete. (Undergraduate
Thesis) De La Salle Univ., Manila, Philippines.
[12] Santos, F., Li, L., Li, Y., & Amini, F. (2011).
Geotechnical Properties of Fly Ash and Soil
Mixtures for Use in Highway Embankments.
World of Coal Ash Conference, Denver, USA.
[13] Toth, S.P., Chan, H.T., & Cragg, C.B. (1987).
Coal Ash as Structural Fill with Special
Reference to Ontario Experience. Can. GeotechJ., 25(1), 694-704.
[14] Kim, B., Prezzi, M., & Salgado, R. (2005).
Geotechnical Properties of Fly and Bottom Ash
Mixtures for Use in Highway Embankments.
Journal of Geotechnical and Geoenvironmental
Engineering ASCE, 131(7), 914-924. doi:
10.1061/(ASCE)1090-0241(2005)131:7(914)[15] Hazelton, P., & Murphy, B. (2007). Interpreting
Soil Test Results: What Do All the Numbers
Mean?. CSIRO, Collingwood VIC.

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DISCRETE ELEMENT MODELLING OF GEOCELL-REINFORCED
SUB-BALLAST SUBJECTED TO CYCLIC LOADING
Ngoc Trung Ngo1, Cholachat Rujikiatkamjorn1 and Buddhima Indraratna1
Centre for Geomechanics and Railway Engineering, Faculty of Engineering and Information Sciences,
University of Wollongong, Wollongong, NSW 2522, Australia. ARC Centre for Excellence for Geotechnical
Science and Engineering, Australia
1

ABSTRACT
Upon repeated train loading, sub-ballast aggregates, placed underneath a ballast layer in rail track, become
degraded and fouled by the progressive accumulation of external fine particles such as mud-pumping of soft
subgrade, seriously decreasing the shear strength and drainage capacity of the track. This paper presents a studyof the load-deformation response of geocell-reinforced sub-ballast under cyclic loads using laboratory tests anddiscrete element method (DEM). A series of large-scale cubical triaxial tests with and without geocell inclusions
are conducted in the laboratory and simulated in DEM to investigate the beneficial effect of the geocells indecreasing the lateral and vertical deformations of railway subballast. Irregularly-shaped particles of sub-ballast
are modelled by connecting and bonding of many circular balls together at appropriate sizes and positions. The
geogcell was simulated by bonding many small spheres together to build a desired geometry and structure. The
load-deformation behaviour of the geocell-stabilised sub-ballast specimen at varied load cycles predicted fromthe DEM modelling agrees well with those measured experimentally, showing that the proposed DEM model inthis study is able to capture the deformation behaviour of the sub-ballast stabilised by the geocell. Additionally,
the DEM modelling also provides insight into the distribution of contact forces, average contact normal andshear forces, which cannot be determined experimentally. These observations clearly prove the reinforcement
effect of the geocell in eliminating the deformation of sub-ballast from a micromechanical perspective.
Keywords: Discrete Element Method, Subballast, Geocell,Cyclic Loading

minimize lateral spreading of the aggregates, hence
maintain stability of the infill granular material [7].
In order to determine the effects of geocells,
different types of geocells have been widely used.
Also, the effects of aperture size and shape andopening area have been investigated by employing
large-scale direct shear box and assessing the shear
strength of unreinforced and reinforced soil [6], [7].
There has been limited research carried out tostudy the effect of geocell mattress on railway
substructure, where the benefits of geocell undercyclic loading have not been investigated in details
either in laboratory or numerical modelling [8]. The
development of computational models that havebeen validated appropriately by either laboratory or
field measurements is thereby inevitable to study theenhanced performance of geocell-reinforced subballast and to derive proper design guidelines forballasted track, considering the confinement effectprovided by geocells [9]. An extensive attempt has
been made in this study to conduct large-scale
cubical triaxial tests of geocell-reinforced sub-ballast
and to develop a discrete element model (DEM)
simulating the composite system, capturing theadditional confinement provided by the geocell tothe infilled sub-ballast.
The discrete element method (DEM) introducedby [10] has been widely used to study stress-strain

INTRODUCTION
In view of rapid urbanization, the demand forsuitable ground improvement solutions is imperative,
in order to construct roads and rail infrastructure
over soft soils. This results in intensified stress onrailway industry to find innovative approaches tomaintain track stability and reduce maintenance cost.
The use of planar geosynthetics (e.g. geogrids,
geotextiles or geocomposites) has proven as apromising technique to strengthen the shear strengthof granular media placed over weak and soft
deposits [1]-[3]. Loss of track geometry that is often
associated with excessive differential settlements
due to localized failure of formation (capping andsubgrade), often leads to decreased stability andreduced track longevity [4]. In this regard, planar
geosynthetics has been effectively utilised to reduce
excessive settlements and lateral displacements ofballasted rail tracks [5]. In recent times, threedimensional cellular reinforcement, also known as
geocell mattress, has been used for different
applications. The improved performance of geocellstabilised soil has been attributed to enhanced
apparent cohesion between the infilled soil and thegeocell [6]. Nevertheless, recent studies have proventhat the additional confinement mobilised during
cyclic loading, helps to enhance confinement and34

GEOMATE- Bangkok, Nov. 14-16, 2016

behaviour of granular materials [11]-[15]. It is notedthat there have been limited research on the use of
DEM to simulate rail sub-ballast under cyclic loads
with high numbers of load cycles. Lu and McDowell[16] carried out DEM analysis to model fresh ballast
under 100 load cycles and showed that the DEM cancapture the stress-strain behaviour of ballast that arecomparable with 500,000 cycles in laboratory. Ngo
et al. [12] conducted DEM analysis to investigate theperformance of geogrid stabilised ballast fouled withcoal, and presented that the interlock of theaggregates with geogrid was the main causes for
improved performance of the composite assembly. It
would also be noted that there has been limited past
DEM studies on the behavior of geocell-reinforced
ballast under a high numbers of load cycles andvaried frequencies. Lobo-Guerrero and Vallejo [17]conducted DEM simulations of model tests for theballast layer subjected to a total of 425 load cycles.
Results indicated that stone-blowing was very
effective while the use of geosynthetics was found to
be less beneficial. This is because the DEM usedcircular bonded particles, and the simulations werelimited to only a few hundred load cycles.
In this study, experimental tests were carried out
and the discrete element method (DEM) was used tomodel geocell-reinforced sub-ballast subjected tocyclic train loading, capturing the deformation andcorresponding micro-mechanical characteristics ofthis composite assembly. The current DEM analysis
was able to include irregular shapes of particles tobetter represent the role of angularity, whereby up to
10,000 load cycles could be performed withdifferent frequencies, and in this respect the current
study is an original attempt to capture the more
realistic behaviour of the plastic deformation of
geocell-reinforced sub-ballast over a much longer
cyclic loading duration.

Fig. 1study.

Cubical triaxial apparatus used in thismm, D min = 0.075 mm, C u = 16.3, C c = 1.3, unit
weight, γ = 18.5 kN/m3). A geocell mattress madefrom polyethylene materials, that was connected at
the joints to create a three-dimensional cellular form
(i.e. depth = 150 mm, ultimate tensile strength = 9.5
kN/m, thickness = 1.3 mm, density = 950 kg/m3)
was used. A predetermined mass of sub-ballast wasplaced inside the cubical box in several layers andcompacted using a vibratory hammer to achieve a
relative density (D R ) of about 77%, which isrepresentative of the density of sub-ballast in thefield [6]. A geocell mattress was placed onto thesurface of the sub-ballast. All specimens wereprepared until the layer of sub-ballast reached a final
height of 450 mm.
The experiments were conducted under planestrain condition, where any lateral movement in thelongitudinal direction (parallel to the track) was
restricted (𝜀𝜀2 =0). The walls were allowed to movelaterally in the direction parallel to the sleeper (ortie) (𝜀𝜀3 ≠ 0), to simulate a long straight section of
track. Laboratory tests were carried out in a stresscontrolled manner, where the magnitudes of thecyclic stresses were determined based on 30
tons/axle load. To investigate the influences ofconfining pressures on the load-deformation of subballast, cyclic tests were conducted at varying
confining pressures of, 𝜎𝜎3 = 5, 10, 15, 20, 30 kPa
and frequencies of 𝑓𝑓 = 10, 20, 30 Hz. Initially, amonotonic strain-controlled load was applied to thespecimen at a rate of 1 mm/min until a mean level ofcyclic deviator stress was attained. Subsequently, astress controlled cyclic loading using a positive fullsine waveform was applied to the specimens where amaximum and minimum stress of q max = 166 kPa
and q min = 41 kPa was used to simulate subballast
under a heavy haul freight network operating in
NSW [7],[18], [19].
Laboratory test results showed that the confiningpressure (𝜎𝜎3 ) and frequency (𝑓𝑓) induce a significant
impact on the load-deformation behaviour of the
sub-ballast under cyclic loading. The experimental
data confirmed that under cyclic loading, geocell
mattress can offer additional confinement (∆𝜎𝜎3 ) to

EXPERIMENTAL STUDY
A large-scale cubical triaxial apparatus (800 mm
long, 600 mm wide and 600 mm high) was designed
and built at the University of Wollongong (Fig.1),
and it was used to investigate the stress-strain
behaviour of the unreinforced and geocell-reinforced
subballast subjected to cyclic loading [6], [9]. The
area of the test specimen in the cubical triaxialchamber was selected based on Australian standard
gauge for heavy haul track with an approximate plan
area of 800mm × 600mm, and 600 mm height. The
sub-ballast material had a total depth of 450 mm, of
which the upper 150mm was stabilised by geocell,as shown in Fig. 1. The material for sub-ballast usedin this study was a locally available crushed basalt,collected from a quarry near Wollongong (NSW,
Australia). The particle size distribution adopted forthe sub-ballast was within the Australian rail
industry specified range (D 50 = 3.3 mm, D max = 19
35

GEOMATE- Bangkok, Nov. 14-16, 2016

qmax

σ1

[𝑨𝑨] [𝑩𝑩]

[𝑨𝑨]

[𝑩𝑩]

𝒌𝒌𝒏𝒏 +𝒌𝒌𝒏𝒏

and the contact shear stiffness is given by:
𝑲𝑲𝑻𝑻 =[𝑨𝑨] [𝑩𝑩]𝒌𝒌𝒔𝒔 𝒌𝒌𝒔𝒔
[𝑨𝑨]

[𝑩𝑩]

75
75

Separate walls to capturinglateral displacement75

σ3Sub-ballast

800

Fig. 2 Schematic DEM model used tocalibrate
sub-ballast
micromechanicalparameters (dimensions in mm).
Table 1 Micromechanical parameters used tosimulate sub-ballast and geocell in DEM
Items
Particle density (kN/m3)
Coefficient of friction
Contact normal stiffness(N/m)
Contact shear stiffness, k s(N/m)
Parameter of contact bondnormal strength, 𝝓𝝓𝒏𝒏 (kN)Parameter of contact bondshear strength, 𝝓𝝓𝒔𝒔 (kN)
Parallel bond radiusmultiplier, r p
Parallel bond normal
stiffness, k np (kPa/m)
Parallel bond shear stiffness,
k sp (kPa/m)
Parallel bond normal
strength, σ np (MPa)
Parallel bond shear strength,
σ sp (MPa)

(3)
(4)

𝒌𝒌𝒔𝒔 +𝒌𝒌𝒔𝒔[𝑨𝑨] [𝑩𝑩] [𝑨𝑨] [𝑩𝑩]where, 𝒌𝒌𝒏𝒏 , 𝒌𝒌𝒏𝒏 ,𝒌𝒌𝒔𝒔 , 𝒌𝒌𝒔𝒔 are the normal stiffness

and shear stiffness of particle A and B, respectively.
The new shear contact force is determined bysumming the old shear force existing at the start of
the time-step with the shear elastic force increment
�𝑭𝑭⃗𝑻𝑻 ← �𝑭𝑭⃗𝑻𝑻 + 𝜹𝜹𝑭𝑭
�⃗𝑻𝑻 ≤ 𝝁𝝁 �𝑭𝑭⃗𝑵𝑵(5)
where, 𝝁𝝁 is the coefficient of friction.

Subballast
15.5
0.72
2.56 E8

Geocell
18.5
0.45
6.51 E6

2.56 E8

6.51 E6

5.36 E6
8.53 E6

43.2
43.2

0.5

0.5

4.86 E7

4.86 ×107

4.86 E7

4.86 ×107

352

352

352

352

DEM Modelling of Cubical Triaxial Test

gradation with a representative field unit weightapproximately of 18.5 kN/m3. Particles weregenerated in the assembly at random orientations toresemble experimental conditions. Micromechanicalparameters to model sub-ballast (e.g. shear andnormal contact stiffness, friction coefficient)selected in the current DEM analysis were
determined based on the process of calibration of
DEM results with the experimental data, aspresented in Table 1. The geocell pocket structure
was modeled by bonding balls of 20 mm-diameterand 10 mm-diameter to form vertical and horizontal

A two-dimension DEM analysis was conducted to
study the interaction between geocell and sub-ballast
by simulating the cubical triaxial tests that werecarried out in the laboratory, as illustrated in Fig. 2.
The angular-shaped grains of sub-ballast were
simulated by connecting a number of circularshaped particles together, mimicking the actual subballast shape and angularity. A total of 26567
particles, with sizes ranging from 0.5 to 19 mm,
were generated to simulate actual sub-ballast
36

450

75

σ3=
5 kPa, 15 kPa

In DEM, the force-displacement law derives thecontact force acting on two particles (e.g A and B) in
contact with the relative displacement between them.
�⃗𝑵𝑵 ) and
The force vector �𝑭𝑭⃗ is described into normal (𝑭𝑭�⃗𝑻𝑻 ) with respect to the contact
shear component (𝑭𝑭plane:
�𝑭𝑭⃗𝑵𝑵 = 𝑲𝑲𝑵𝑵 ∙ 𝑼𝑼𝒏𝒏
(1)�⃗𝑻𝑻 = −𝑲𝑲𝑻𝑻 ∙ 𝜹𝜹𝑼𝑼𝒔𝒔
𝜹𝜹𝑭𝑭(2)
where, 𝑲𝑲𝑵𝑵 and 𝑲𝑲𝑻𝑻 are the normal and shear
stiffnesses at the contact; 𝜹𝜹𝑼𝑼𝒔𝒔 is the incremental�⃗𝑻𝑻 is the incremental
shear displacement, and 𝜹𝜹𝑭𝑭
shear force.
The normal contact stiffness for the linear contact
model used in this study was computed as:
𝒌𝒌𝒏𝒏 𝒌𝒌𝒏𝒏

75

σ3=5 Kpa, 15 kPa

DISCRETE ELEMENT MODELLING𝑲𝑲𝑵𝑵 =

Time

qmin

75

the infill material (i.e. other than the confiningpressure available from sleepers and shoulder
ballast), and help to decrease the axial and lateral
deformations [9]. Also it is noted that due to cyclicloading, the magnitude of would be increase as thenumber of load cycle increases and when thedensely compacted infill material dilates and thereby
increases the magnitude of hoop stress. A summaryof the key aspects of experimental outcomes can be
brieftly summarised as: (i) the mobilised hoop stressof the geocell pockets is generated as a result of thedilation of the infilled soil during shearing; and (ii)the magnitude of hoop stress varies with the geocell
modulus [6], [7], ],[18].

GEOMATE- Bangkok, Nov. 14-16, 2016

80

panels, respectively. This simplified geocell
structure was presumed to be adequate to provide
the confinement effect for the sub-ballast packed
inside the cellular pockets. Micromechanicalparameters (Table 1) to model the geocell were
determined based on a series of simulated tensile
tests and compared the tensile force-strain responsewith data measured experimentally.

10Hz
20Hz
30Hz

Unstablezone

60

Settlement (mm)Lab. DEM

Without Geocell

70

With
Geocell

Unstable
zone

50
40
30
20
10

RESULTS AND DISCUSSION

00

The DEM model for the plane strain cubicaltriaxial tests was used to simulate geocell-reinforced
sub-ballast subjected to a confining pressure of
𝜎𝜎3 =10 kPa and cyclic frequencies of 10, 20, and 30
Hz, similar to the loading conditions conducted in
the laboratory. DEM simulations to model the subballast with and without geocell inclusions were
simulated up to 10,000 load cycles, where most of
the sub-ballast deformation had taken place as
observed in the laboratory. During loading, vertical
positions and lateral movements of the sub-ballast
assemblies were recorded to determine theassociated settlements and lateral displacements atcorresponding load cycles.

2000 4000 6000 8000 10000

No. of Cycles , N

Fig. 302000 4000 6000 8000 10000

No. of Cycles , N

Settlement versus load cycles measured
experimentally and predicted in DEM(modified after Ngo et al. 2015).

ballast specimen subjected to the cyclic load at agiven frequency of 20 Hz at a settlement (S) of 5mm, while Figs. 4b-d show the contact force
distributions of geocell-reinforced sub-ballast at
settlements of S=5 mm, 15 mm, and 20 mm,
respectively. It is noted that the contact forcesamong particles are plotted as lines on the same
scale, whose thickness is proportional to its
magnitude, and for clarity, only those contacts witha magnitude exceeding the average force of thewhole assembly are presented. They clearly showthat the total number of contact forces and maximum
contact forces increase as settlement increases, andthis can be attributed to the assembly was compacted
and compressed to sustain the external load. For
instance, with reinforced sub-ballast, the number ofcontacts is 60,252 for the settlement of 5 mm, and it

Settlements of Sub-ballast with and without
Geocell
The average accumulated settlement at different
load cycles obtained from DEM simulations
compared to the experimental results are presented
in Fig. 3. Results obtained from DEM analysis
matched reasonably well with the experimental dataat any given frequency and confining pressure. The
predicted and measured data indicated that thesettlement increased with an increase in frequency.
Geocell-reinforced sub-ballast exhibited lesssettlement than that of the unreinforced assembly.
Undoubtedly, this is a result of additional confiningpressure provided by the geocell would decrease thedeformation of sub-ballast. When the sub-ballast
aggregates were compacted over a geocell, they
were projected through the geocell pockets andgenerated a strong mechanical (i.e. acting as a nondisplacement boundary) which results in reduced
settlement. Additionally, the settlement accelerated
significantly during the first few thousand cycles due
to initial particle compression and rearrangement,
and then the settlement increased at a diminishedrate in the subsequent load cycles and approached an
approximately constant rate at very high load cycle.
Contact Force Distributions of Geocell-reinforced
Sub-ballast
Contact forces in a sub-ballast assembly aretransferred through an inter-connected network offorce chains via contact points. Fig. 4a presents
contact force distributions of an unreinforced sub-Fig. 4
Distribution of contact forces for
unreinforced and reinforced subballast at varied
settlements (modified after Ngo et al. 2015)37

GEOMATE- Bangkok, Nov. 14-16, 2016

increases to 78,252 and 83,521 contacts forsettlements of S=15 and 20 mm, respectively.
Maximum contact forces also increases with anincrease of settlements, and these are 745 N, 857 N
and 946 N for the settlements of S=5, 15 and 20 mm,
respectively. Compared to the unreinforced subballast (Fig. 3a), the reinforced assemblies createdmore contacts within the geocell regions, and this
could be associate
d with the confinement thegeocell provided to the infilled sub-ballast. It canalso be seen that the tensile forces (in red colour) in
geocells are mobilised with an increased insettlement.

parameters to simulate sub-ballast and geocell were
determined by comparing with laboratory test data.

Variations of Contact Normal and Shear Forces
Fig. 5 presents the variations of contact normal andshear forces with depth for reinforced andunreinforced sub-ballast (at N=10,000 cycles).
Compared to the unreinforced cases (Figs. 5c, 5d),geocell-reinforced sub-ballast assemblies exhibit asignificant increase in the contact force within thegeocell zone, but underneath the geocell the average
normal and shear contact forces decrease with depth
and approach almost constant values near the bottomof the assembly. Undoubtedly the inclusion of
geocell decreases the shear and normal contact force
in sub-ballast below the geocell. It is worth
mentioning that the micro-mechanics of the geocellreinforced sub-ballast conducted in this study was
limited to the distribution of contact force chains andthe average contact normal and shear force
distributions. However, the comparison of theexperimental observations with the 2D plane strain
DEM analysis proves that to the current analysis wasable to capture the load-deformation behaviour of
geocell-stabilised sub-ballast in spite of theselimitations. Generally, the authors have made a few
simplifications to keep the micro-mechanicalanalysis fairly simples, as the requirements ofbrevity of this paper would not allow the reportingof more detailed DEM analyses that could capture
other micro-mechanical aspects such as theevolutionof fabric anisotropy and complex
detailing of changing angularity with the highnumber of loading cycles.

Fig. 5 Distributions of contact normal and shear
forces: (a) and (b) - with geocell inclusion; (c)and (d) - without geocell.
Once these parameters were properly validated, they
were used to simulate the cubical triaxial tests fortesting sub-ballast subjected to cyclic loading at
frequencies of 10 Hz, 20 Hz and 30 Hz.
Experimental data of settlements and lateraldisplacements were comparable with those obtained
from DEM simulations at a given frequency andconfining pressure, indicating that the DEM model
proposed in this study could simulate the loaddeformation behaviour of a geocell-reinforced subballast assembly. As the frequency increased, thesettlement and lateral deformations of sub-ballast
increased, but unlike the unreinforced sample, thegeocell-reinforced sub-ballast exhibited remarkably
less deformation. This was undoubtedly attributed tothe confinement provided by geocell that prevented
sub-ballast aggregates from free movement thatwould otherwise occur.
Contact force distributions of geocell-reinforced
sub-ballast were presented. DEM results showed thatthe total number of contact force distributions andthe maximum contact force increased with increased
deformation. The contact normal and shear forcesdeveloped among sub-ballast particles at varied
depths were also captured. The magnitudes of theseforces within the geocell zone were considerably
higher than at other locations. Underneath thegeocell, these contact forces continuously decreased

CONCLUSION
A series of large-scale cubical triaxial tests werecarried out on sub-ballast with and without geocell
inclusion, and then the results were used to calibrate
and compare with the DEM analysis. Irregularparticles of sub-ballast were simulated by clumping
several circular balls together to represent
appropriate angularity. Geocell was modelled in 2D
plane strain DEM model by bonding small balls
together to form the cellular pockets with contact
and parallel bonds. A set of micromechanical38

GEOMATE- Bangkok, Nov. 14-16, 2016

with depth and approached almost constant values
near the bottom of the granular assembly.

[9] Ngo, N. T., Indraratna, B., Rujikiatkamjorn, C. ,and Biabani, M. (2015). "Experimental anddiscrete element modeling of geocell-stabilized
subballast subjected to cyclic Loading”, J.
Geotech. Geoenviron. Eng.,42(4):04015100.
[10] Cundall, P. A. , and Strack, O. D. L. (1979). "A
discrete numerical model for granularassemblies". Géotechnique. 29(1): 47-65.
[11] Lim, W. L. , and McDowell, G. R. (2005).
"Discrete element modelling of railway ballast".
Granular Matter. 7(1): 19-29.
[12] Ngo, N. T., Indraratna, B. , and Rujikiatkamjorn,
C. (2014). "DEM simulation of the behaviour ofgeogrid stabilised ballast fouled with coal".
Computers and Geotechnics. 55: 224-231.
[13] Indraratna, B., Ngo, N. T., Rujikiatkamjorn, C. ,and Vinod, J. (2014). "Behaviour of fresh andfouled railway ballast subjected to direct shear
testing - a discrete element simulation". Int. J.
Geomech. , 14(1): 34-44.
[14] Indraratna, B., Ngo, N. T., Rujikiatkamjorn, C.
& Sloan, S. W. (2015). Coupled discreteelement–finite difference method for analysing
the load-deformation behaviour of a single stone
column in soft soil. Computers and Geotechnics,
63 267-278.
[15] Ngo, N. T., Indraratna, B. & Rujikiatkamjorn, C.
(2015). A study of the behaviour of fresh andcoal fouled ballast reinforced by geogrid using
the discrete element method. In K. Soga, K.
Kumar, G. Biscontin & M. Kuo (Eds.),
Geomechanics from Micro to Macro (pp. 559563). United Kingdom: Taylor & Francis. 2014.
[16] Lu, M. , and McDowell, G. R. (2010). "Discreteelement modelling of railway ballast undermonotonic and cyclic triaxial loading".
Geotechnique. 60(6): 459-467.
[17] Lobo-Guerrero, S. , and Vallejo, L. E. (2010).
"Crushing of particles under simulated staticand centrifuge forces". GeoFlorida 2010:
Advances in Analysis, Modeling & Design(Geotechnical Special Publication, 199), ASCE.
550-559.
[18] Indraratna, B., Biabani, M.M., Nimbalkar, S.,
2015. Behaviour of geocell reinforced
subballast subjected to cyclic loading in planestrain condition. J. Geotech. Geoenviron. Eng.
141 (1), 04014081-1-16.
[19] Ngo, N. T, and Tung, T. (2016). “Coupleddiscrete-continuum method for studying loaddeformation of a stone column reinforces rail
track embankments”. Procedia Engineering, 142
138-144.

ACKNOWLEDGEMENTS
The Authors would like to thank Australian
Research Council (ARC) for providing funding tothis study through the ARC Centre of Excellence inGeotechnical Science and Centre for Geomechanics& Railway Engineering (CGRE), University of
Wollongong. The Authors thank to Dr Mahdi
Biabani and laboratory assistance from technicians,
Alan Grant and Ian Bridge during the laboratory
tests. The Authors would also appreciate the kindpermission from the Journal of Geotechnical andGeoenvironmental Engineering - ASCE forreproducing selected content in this paper.
REFERENCES[1] Bergado, D. T., Chai, J. C., Abiera, H. O.,
Alfaro, M. C., and Balasubramaniam, A. S.
(1993). "Interaction between cohesive-frictionalsoil and various grid reinforcements". Geotext
Geomembranes. 12(4): 327-349.
[2] Indraratna, B., Ngo, N. T., and Rujikiatkamjorn,
C. (2011b). "Behaviour of geogrid-reinforced
ballast under various levels of fouling". Geotext
and Geomembranes. 29(3): 313-322.
[3] Tutumluer, E., Huang, H. , and Bian, X. (2012).
"Geogrid-aggregateinterlockmechanism
investigated through aggregate imaging-based
discrete element modeling approach". Int J
Geomech. 12(4): 391-398.
[4] Selig, E. T. , and Waters, J. M. (1994). “Track
geotechnology and substructure management”,
Thomas Telford, London.
[5] Indraratna, B., Ngo, N. T. , and Rujikiatkamjorn,
C. (2013). "Studying the deformation of coalfouled ballast stabilised with geogrid undercyclic load". J. Geotech. Geoenviron. Eng.,
139(8): 1275–1289.
[6] Biabani, M. M., Indraratna, B., and Ngo, N. T.
(2016). “Modelling of geocell-reinforced
subballast subjected to cyclic loading”. Geotext
Geomembranes. 44(4), pp: 489-503.
[7] Biabani, M. M., Ngo, N. T., and Indraratna, B.
(2016). “Performance evaluation of railway
subballast stabilised with geocell based on pullout testing”. Geotext Geomembranes.44(4):579–591.
[8] Leshchinsky, B. and Ling, H., 2013a. Effects ofgeocell confinement on strength anddeformation behavior of gravel. J. Geotech.
Geoenviron. Eng., 139(2), 340-352.

39

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ANALYSIS OF DCM WALL SUPPORTED EXCAVATION IN CLAY BY
GROUND SPRING MODEL

1,2

Siriwan Waichita1, Pornkasem Jongpradist2 and Chanchai Submaneewong3
Faculty of Engineering, King Mongkut's University of Technology Thonburi, Thailand;
3
Soilcrete Technology Co., Ltd., Thailand
ABSTRACT

Nowadays, an application of Deep Cement Mixing (DCM) as a retaining structure in deep excavation works
becomes gradually popular in very dense population areas because the construction process causes less noise andenvironmental impact than other systems. In addition, this new kind of retaining structure has various forms ofutilization that depend on the designer’s experience and considerations. For better understanding of this system,
full scaled test, down scaled physical model test, and numerical analysis are required to tackle the interested
problem. To effectively discuss the behaviors observed from the full-scale numerical analysis and physical model
test, the scaling factor must be seriously considered in the model test. However, it is difficult to scale down theproperties of soft clay in the physical model test. Therefore, the soil and its lateral pressure transferred to the wall
are modeled as series of springs and lateral forces in the model, respectively. To ensure the effectiveness of thismodeling, preliminary evaluation is necessary. In this study, a 2D plane strain Finite Element model of an
excavation with the DCM wall as a retaining structure was compared to a 3D Finite Element model with a seriesof ground springs to take the lateral stiffness of the in-situ soil behind the wall into consideration. The results ofthis numerical investigation reveal that the ground spring model has sufficient accuracy to represent the lateralsoil-structure interaction.
Keywords: Deep Cement Mixing, Retaining Structure, Excavation, Ground-springon DCM wall behavior. To fulfill this lack ofknowledge, a series of numerical analysis, physical

INTRODUCTION
Thick layer of soft soil was found in many wellknown areas of the world, e.g., Thailand, Japan, andSweden. Since the demand of land utilization in thecity becomes increasing with time, large number of
new buildings are commonly constructed withbasement levels. During the excavation process,problem relevant to lateral movement is expectedly
occurred. To control the excessive displacement thatwill affect to the adjacent structures, retaining wallis needed. The commonly used walls are either steel
sheet pile, diaphragm wall, or contiguous pile wall.
Some of congested urban areas require extra
conditions of construction method such as very low
noise and vibrating [1]. Deep Cement Mixing wall,as an alternative of retaining structure is introduced
to meet the requirements. There were several reportsof case history using Deep Cement Mixing wall to
support the excavation in various forms ofapplications, e.g., DCM wall without bracing [1],
DCM wall with wall-strut [2], combination of DCM,
sheet pile, and tie back supported excavation [3], andDCM cross walls installation with diaphragm wall
[4] and shown in Fig. 1. Utilization of DCM in
excavation were not only the support system, butalso a ground improvement for soft soil in passivezone of an excavation [5]. However, they are still
highly empirical in terms of analysis and design withseveral assumptions due to unclear understandings

model test, and full-scale test are important tosystematically tackle the problem. Despite that the
full-scale test is the most reliable method to study theactual behaviors because the actual condition can be
reproduced. The test expense is high and it isdifficult to repeat the test with constant condition.
While physical model requires the complicatedscaling down technique, it can control thecircumstances and the interested parameters can bevaried. To compensate the limitations of full-scaled
test, the study on DCM wall thus focuses on physicalmodel test and numerical analysis.
In order to scale down the problem from the field
for setting up the physical model in the laboratory,
scaling law is the significant factor in theconsideration. Because the test is to conducted under1g condition, the properties of soft clay surrounding
the wall cannot be correctly scaled down to meet therequired values during the preparation if large
scaling factor is considered. To solve this problem,
set of springs is introduced to substitute the soil in
the unexcavated side whereas point loads are
represented as lateral loads of soil on the excavatedside. Consequently, only the small scale DCM wall
is to be prepared in the test. However, thecontinuously distributed pressures (both excavation
and un-excavation sides) are discretely represented
40

GEOMATE- Bangkok, Nov. 14-16, 2016

by numbers of springs and point loads in the physical± 0.00± 0.00± 0.00

-1.00
16.45
-5.70

11.00

DCM
Wall

14.00

DCM wallDiaphragm
wall-10.00

33.50
-7.00

-20.0040.00

Sheet Pile
DCM Wall

Wall-strut
orCross wall

Wall-strut

2.90a) DCM wall without bracing [1]Fig. 1

0.76

3.10

1.00

b) DCM wall with wall-strut [2] c) DCM with tie-back sheet pile wall [3] d) Diaphragm wall with DCM Cross wall [4]Various patterns of DCM wall used in case histories.

model tests. It is thus necessary to evaluate the effectof numbers of springs and point loads on the model
accuracy. Moreover, the minimum require numbersof both springs and point loads are preferred tominimize the effort during the preparation and test.
To achieve this, preliminary analysis of DCM wall
excavation using ground spring model in comparison
to conventional continuum mechanics is carried out
in this study. In the analyses with ground springmodel, numbers of springs and point loads are varied
and the results are discussed with those from thefinite element analysis on the basis of same
excavation problem.

cement columns were overlapped for the 2.8 m wide
DCM block pattern wall with a depth of 15 m. The
walls were installed by a jet grouting method with acement content of 250 kg/m3 of soil to obtain thedesigned unconfined compressive strength of 1200
kPa at curing time of 28 days. During theconstruction process, horizontal movements alongwall depth were observed by inclinometer forconstruction control and the validation.
Due to the symmetrical geometry of the
excavation, half of the problem was modeled inplane strain condition using PLAXIS 2D software asshown in Fig. 2. The soil parameters were obtained
from experimental and empirical data, listed on
Table 1. Since the excavation was completed within
less than 3 months, undrained analysis was used in
the simulation. A retaining wall was classified as asmall strain characteristic structure [6]. Numericalsimulation of various cases of deep excavation in
Bangkok subsoil revealed that the excavation
simulation by using the Hardening soil model withsmall strain (HSS) provided high accurate results [7].
The HSS was thus adopted to represent the soft tostiff clay behaviors, while Mohr Coulomb model(MCM) was selected for fill crust, clayey sand and
DCM wall. The excavation was simulated following
the actual construction sequence. Figure 3 showed
the comparison between analysis result and fieldobservation of wall horizontal displacements

2D NUMERICAL ANALYSIS OF THE
REFERENCE CASE STUDY
The excavation with un-braced DCM wall was
adopted to validate 2D numerical model used as areference in this study. This excavation wasconstructed in Bangkok subsoil as shown in Fig. thatthe soil stratum consists of 2.5 m thick fill crust
overlying 13.5 m thick soft clay. Stiff clay layer
started at a level of -16.00 m with the thickness of 5m. There is 1 m of clayey sand, sandwiched between
the stiff clay layer and 12 m thick very stiff clay. The
excavation was conducted to the maximum
excavation depth and width of 5 m and 27 m,
respectively. Three rows of 1 m in diameter deep
Table 1 Materials Properties
Materials

Model

c'



E50ref

Eu
MPa

ref
Eoed

Eurref

Gmax

 0.7

ur 

MPa
MPa
MPa
MPa
kPa
deg
DCM
MC
cu=6000300
Fill Crust
MC
cu=60
0
42
Very Soft Clay
HSS
1
21
7.2
7.2
36
72
1x10-3
0.2
Soft Clay
HSS
1
22
12.2
12.2
48
75
8x10-40.2-5
Stiff Clay
HSS
15
26
45
45
150
170
1x10
0.2
E ' =80
Clayey SandMC
036
Very Stiff Clay
HSS
40
26
80
80
300
350
1x10-50.2
Remarks: MC was stand for Mohr Coulomb Model, and HSS was Hardening Soil Model with Small Strain.
pref =100 kPa, and m=1for all materials using HSS.
41

GEOMATE- Bangkok, Nov. 14-16, 2016

2.80

13.50

Fill Crust

2.50

5.00

Very Soft Clay

6.50

DCM Wall

7.00

Soft Clay

5.00

Stiff Clay

1.00

Clayey Sand

12.00

Very Stiff Clay

15.00

43.50

Fig. 2

Geometry and mesh of 2D plane strain model.
0

influence parameters. Therefore, physical model test
is preferable to overcome these limitations. In casethat the test will be carried out in the laboratory
under 1g condition, the scale down technique must
be involved. Modeling of the studied problem iscomposed of DCM wall, surrounding clay layer, andearth pressure. Due to the fact that the height of wallin physical model is limited by the ceiling of theroom and device capacity, the scaling factor must belarge enough to scale down the actual 15 m high wallin laboratory test. The difficulties in the preparationof correctly scaled down clay layer thus occur. An
idea of using set of springs and point loads instead ofclay both sides of the scaled down wall was
introduced and it is called as “ground spring model”.
The set of springs and point loads are discretely
applied on the wall. To do a feasibility study andensure that this model can represent the excavation
work, therefore the preliminary analysis of using theidea of ground spring model in excavation is
performed.
The validated 2D model of field case study andinput parameters in previous section were used as
references in this analysis. ABAQUS program wasutilized for a three dimensional analysis of a unit cell
excavation. Mohr Coulomb Plasticity was employedfor DCM wall properties. To reduce complexities ofthe problem, the soil layers are simplified ashomogeneous layer in both of 2D and 3D models.
Set of point loads was calculated from earth pressureof soil, which were varied in each excavation
sequence. In the same manner, stiffness of spring in
Eq. (1) is depended on horizontal subgrade reaction
value and interval length. Vesic’s equation [8] was

Depth (m)

5

10

15

Field -5.00m
Analysis -5.00m
20

0

5

10

15

20

25

30

35

40

Horizontal Displacement (mm)Fig. 3Comparion of horizontal displacement fromfield observation and analysis.

occurred at the final level of excavation. It is seenfrom the figure that good prediction was provided in
the validation, the analysis by 2D plane strain model
together with the parameters used can reasonably use
as a reference model.
3D NUMERICAL ANALYSIS WITH GROUNDSPRING MODEL
Although field case study is the most reliable
method to study the structural behaviors, it cannot
provide the failure state of excavation and vary the42

GEOMATE- Bangkok, Nov. 14-16, 2016

5.00

1.25

1.25

1.67

1.25

1.67

2.00

Fig. 4(a)(b)(c)
Comparison of models in parametric study of spring quantities; (a) 8 springs case, (b) 10 springs case,and (c) 12 springs case.

was adopted for horizontal subgrade reaction
calculation as shown in Eq. (2). Basis of horizontal
subgrade reaction was applied in laterally loaded
piles [9] and excavation [10].

K h  ksh  B  L

0.65Es Es B 412ksh 
B(1  2 ) Ew I w
Which

Excavation Side

Point Load applied
to a rigid plate(1)

Spring tied toa rigid plate

15.00

(2)

: Horizontal stiffness of spring: Horizontal subgrade reaction: Width of wall
: Interval length of spring: Soil’s elastic modulus: Poisson’s ratio of soil
: Wall’s elastic modulus: Wall’s moment of inertia

60

0.

Kh
ksh
B
L
Es
ν
Ew
Iw

Un-excavation Side
(Ground)2.80

Fig. 5Assembled model and mesh of 3D withground spring model.

RESULTS

The purposes of this analysis are not only to checkthe reliability of spring model but also to be a
prototype in the down-scaled physical model test.
Parametric study of spring quantities was thus
conducted. Figure 4 shows a comparison of eachcase in the parametric study including modeling with
8 springs, 10 springs, and 12 springs. As the wall
depth is kept constant, an interval length of springs
is varied by number of springs. Three different
interval lengths of springs including 1.25 m, 1.67 m,
and 2 m, was considered in the models. Rigid plate
was used to improve the interaction between springs
or loads to DCM wall. Assembled model was shownin Fig. 5, composing of rigid plates tied with springsin the ground (unexcavation) side and rigid plates
tied with point loads in the excavation side.
Following to the construction sequences, each spring
was removed to simulate an excavation of soil layer.
The final level of excavation was located at 5 m fromwall top.

In this section, the analysis results from groundspring model with different numbers of springs andpoint loads are compared together with that of theresult from 2D plane strain continuum analysis. The
results to be shown and discussed include horizontal
displacement, principal and shear stresses. It is noted
again that all analyses consider the same problem
under homogeneous soil layer.
Horizontal displacement, a common parameterused in excavation monitoring, was shown in Fig. 6.
It is seen from the figure that all cases with groundspring model represent a good tendency with thereference 2D continuum analysis. Not only the shapeof horizontal displacement profile that are resemble,
the predicted values are also in the same order withresults from 2D analysis. The maximum horizontal
displacement occurred at the top and decreased withdepth. The inflection point of the horizontal
displacement profile of all cases appear at the depth43GEOMATE- Bangkok, Nov. 14-16, 2016of about 5 m which is the excavation level. Among
three cases of ground-spring analyses, the horizontal
displacement decreases with increasing number ofsprings. However, only drastic increase of horizontal
displacement was observed when spring quantity
was changed from 8 to 10. In contrast, insignificant
change can be seen in the region below the
excavation level when the number of springs
increased from 10 to 12. This implies that 10 springs
are sufficient, for this case study, to simulate theDCM wall excavation.

as concrete lean, are utilized to suppress this adverse
effect. Reanalysis of 2D continuum model with
applying the overburden pressure back to the
excavated level was performed and the comparison
was shown in Fig. 11. When soil heave waseliminated, the maximum shear stress decreases tothe same way of the 3D model results. The
investigation prove that, if the effect of soil heave
was eliminated or sufficiently minimized, the groundspring model can capture the shear stress distribution
in the DCM wall.

0

0

2

2
4

4

Excavation Level -5.00m
Depth (m)6Depth (m)Excavation Level -5.00m
DCM wall6

DCM wall

8

8
10

10

Observation line
2D-Plane Strain
3D-8 Springs
3D-10 Springs
3D-12Springs

12
Observation line

1214

2D-PlaneStrain
3D-8Springs
3D-10Springs
3D-12Springs

14

16
-200

-150

-100

-50

0

50

100

Major Principal Stress (kPa)

160Fig. 6

10

20

30

40

Fig. 7

Major principal stress at center of the wall.

Horizontal Displacement (mm)0Horizontal displacement results.

2
4

Considering at the center of wall thickness, thepredicted major and minor principal stresses fromthree analysis cases by ground-spring model gavethe same tendency with continuum analysis results
as shown in Figs. 7-8. All analysis cases yield similar
results, particularly for the depth located above thefinal excavation level. The results between 8 and 10
springs cases for the depth located below the
excavation level was slightly different but almost thesame value was obtained when increasing thenumber of spring from 10 to 12. This observation
also suggests that the minimum required number ofspring is 10. The maximum shear stress distribution
along the wall height is shown in Fig. 9. The
predicted results from all cases have good agreement
to that from continuum analysis except those in therange of 5 m below excavation level. For the depth
above excavation level, all cases gave nearly thesame value. The results below the excavation level
were much different from that of the continuum
analysis. The predicted results by soil-spring model
between using 10 and 12 springs were very close.
Since the excavation process reduces overburdenpressure, vertical displacement results of 2D planestrain analysis (Fig. 10) showed that soil heave
occurred at the excavated side of the wall. Note that,the 3D model with soil-spring does not consider thiseffect. In real practice, some counter measures, such

Excavation Level -5.00m
DCM wall

Depth (m)6
8
10

Observation line
2D-Plane Strain
3D-8 Springs
3D-10 Springs
3D-12Springs

12
14
16
-300

-250

-200

-150

-100

-50

0

50

Minor Principal Stress (kPa)

Fig. 8

Minor principal stress at center of the wall.
0
2D-Plane Strain
3D-8 Springs
3D-10 Springs
3D-12 Springs

2
4

Excavation Level -5.00m

Depth (m)6
8
DCM wall1012
14

Observation line
16
0

20

40

60

80

100

Maximum Shear Stress (kPa)

Fig. 944

Maximum shear stress at center of the wall.

GEOMATE- Bangkok, Nov. 14-16, 20168

excavation bottom occurred. Under the reference
excavation case, the minimum required number ofsprings is 10.

Vertical Displacement (mm)76ACKNOWLEDGEMENTS

5

The authors gratefully acknowledge support
from Research and Researchers for Industries (RRi)and The Thailand Research Fund (TRF) undercontract grant PHD57I0076.

4
DCM wall3

Observation line

2
1

REFERENCES
0

2

4

6
8
Distance (m)

10

12

14

[1] Tanseng P, “Soil-cement column wall without
Bracing for Mat Foundation Construction in
Bangkok Sub-Soils”, in Proc. 14th AsianRegional Conference on Soil Mechanics andGeotechnical Engineering, 2011.
[2] Tanseng P, “Soil-cement column wall withwall-strut to minimize ground movement for aroad tunnel construction in Bangkok subsoil”, in
Proc. 38th Int. Conf. At World Tunnel Congress,
2012.
[3] Yang DS, “Soil-Cement Walls for Excavation
Support”, in Proc. Int. Conf. On Earth Retention
Systems, 2003.
[4] Ou CY, Hsieh PG, and Lin YL, “A ParametricStudy of Wall deflections in Deep Excavations
with the installation of cross walls”, Computers
and Geotechnics, Vol.50, pp. 55-65.
[5] Pavan A, and Tamilmani T, “Numerical
Analysis on the Effect of Jet Grout Piles on an
Excavation Located in an Urban Area”, Int. J. of
GEOMATE, Vol.8, No.1, March 2015, pp.
1167-1171.
[6] Atkinson JH, and Sallfors G, “Experimental
determination of stress – strain – time
characteristics in laboratory and in situ tests”, in
Proc. 10th European Conf. On Soil Mechanics
and Foundation Engineering, 1991, pp. 915-956.
[7] Rukdeechuai T, et al., “Influence of Soil Models
on Numerical Simulation of Geotechnical works
in Bangkok Subsoil”, Journal of Research in
Engineering and Technology, Vol.3, 2009.
[8] Vesic AS, “Beam on elastic subgrade and thewinkler hypothesis”, in Proc. 5th Int. Conf. Soil
Mechanics and Foundation Engineering, 1961.
[9] Bahrami A, and Nikraz H, “ Initial Soil Spring
Stiffness for Laterally Loaded Piles” in Proc.
11th Australia New Zealand Conference onGeomechanics, 2012.
[10] Ou CY, Deep Excavation Theory and Practice.
London: Taylor & Francis Group, 2006, ch. 7.

Fig. 10 Vertical displacement along excavation
width from 2D continuum analysis.
0
2
4

Excavation Level -5.00m

Depth (m)68
102D-Plane Strain
2D-Plane Strain-Anti Heave
3D-10 Springs

12
14
16
0

20

40

60

80

100

Maximum Shear Stress (kPa)

Fig. 11 Maximum shear stress at center of the walland effect of soil heave.
CONCLUSION
A case study of Deep Cement Mixing wall as aretaining structure in deep excavation work in
Bangkok subsoil was numerically simulated under
2D plane strain assumption and 3D approach using
ground-spring model. The 2D simulation was
validated by field observation using the monitored
horizontal displacement of the wall. Then the 2D
simulation can be used as a reference case to verify
the proposed analysis method using the ground spring model. The model is composed of a set ofsprings representing the ground in front of the wall
(excavated side) and a series of point loads
representing the lateral ground pressure applied to
the wall (unexcavated side). Analyses by varying
number of springs were carried out and both stressand deformation were observed in the study.
According to the investigated results, the groundspring model can reasonably reproduce the stressedinduced in the wall as well as wall deflection
behavior. However, the reproduced stresses in thewall can be much altered if the soil heave at the45

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

PROBABILISTIC SAFETY ASSESSMENT OF ANCHORED SLOPE
Takashi Hara1, Akira Tomioka2, Nobuyuki Soga3 and Atsushi Yashima4
Department of Civil Engineering, Gifu University, Japan; 2-4 Gifu University, Japan

1

ABSTRACTThe maintenance of existing ground-anchored slopes has been traditionally carried out by experts’ empirical
judgment based on tensile force measurements of partial anchors composing the target-anchored slope. The
experts’ judgment is, however, more or less qualitative, so it is difficult to apply former maintenance experiences
to other slope maintenance practices. Moreover, because the judgment is made through a technical committee,
the present maintenance scheme takes long time and maintenance engineers cannot maintain the slope
themselves rapidly. If the former maintenance experiences based on the judgments are organized quantitatively,
rapid and reasonable maintenance of the slopes by maintenance engineers will be able to be achieved.
Therefore, the authors have conducted a study on development of a probabilistic safety assessment approachfor existing ground-anchored slopes, which can reproduce the experts’ judgments applied to former
maintenances quantitatively. As the result, practicability of the proposed probabilistic safety assessment
approach on the present stage of this study could be confirmed. This paper describes the practicability of the
proposed approach from comparisons between the safety evaluation obtained from the proposed approach andmade from experts’ judgments with respect to actual existing anchored slopes as well as the detail of the
proposed probabilistic safety assessment approach.
Keywords: Ground-Anchor, Existing Slope, Maintenance, Probabilistic Safety Assessment (PSA)

Therefore, the authors have conducted a study on
development of a probabilistic safety assessment
approach for existing ground-anchored slopes,
which can reproduce the former maintenanceexperiences based on experts’ empirical safety
judgment quantitatively. This paper describes a
proposed probabilistic safety assessment model foran anchored slope and its reproducibility of thepresent maintenance from several examples.

INTRODUCTION
Tensile forces acting in existing ground-anchors
maintaining slope stability vary depending on thepassage of time after the installation. The variation
degree of the tensile forces with respect to design
one differs due to several phenomena occurred ineach site of existing slopes, such as creep of PC
strand and/or slope ground, partial deformation of
the slope, and so on. Furthermore, the existing
excessive tensioned anchors let us know dangerous
situation of the anchored slope. The danger level isdifferent depending on the tension intensity, thedistribution of the anchors with high tensile force
and so on. Therefore, the safety judgment of existing
ground-anchored slopes is difficult. Hence, thepresent maintenance has been traditionally carriedout with experts’ empirical safety judgment based on
tensile force measurements of partial anchors
composing of an entire existing anchored slope in atechnical committee. The experts’ empirical
judgment is, however, qualitative, thus theapplication of accumulated former maintenanceexperiences to other slope maintenances is difficult.
Moreover, because the judgment is made through atechnical committee, the present maintenance takes
long time and maintenance engineers cannot
maintain the slopes themselves rapidly. If former
maintenance experiences based on the judgments areaccumulated quantitatively, rapid and reasonable
maintenance of the slopes by maintenance engineers
will be able to be achieved.

AIM OF THIS STUDY
Fig. 1 shows the present maintenance procedureof an existing ground-anchored slope in Japan.

Fig. 1 Present maintenance procedure.
46

GEOMATE- Bangkok, Nov. 14-16, 2016

In the present procedure, the tensile force acting
in partial ground-anchors, 5 - 10% of the entire
target slope’s ones, passed 10 - 20 years from theinstallation are measured and the safety level of theanchored slope is judged by experts in a technicalcommittee qualitatively. In the safety judgment, thetreatment is chosen from following ones; (1)observation, (2) additional survey, and (3)countermeasure execution. In additional survey, thetarget surveys are selected from following ones; (1)additional boring, (2) additional measurement oftensile forces acting anchors, (3) measurement ofslope movement and etc., as well as their execution
points and volumes are planned. Then the safety ofthe target slope is reevaluated based on the results of
the additional surveys.
On the other hand, the present maintenance has
several problems. The application of theaccumulated former maintenance experiences toother existing slope maintenances is difficult
because the experts’ empirical judgment is more orless qualitative. The technical committee meetings
for making the experts’ empirical judgment are held
three or four times a year, so the present
maintenance takes long time and maintenanceengineers cannot maintain the target slope
themselves rapidly.
The authors, in order to solve these problems,
have conducted a study on development of aprobabilistic safety assessment approach for existing
ground-anchored slopes, which can reproduce theformer maintenance experiences of existing slopes
quantitatively. Namely, the target of this study is to
propose a future quantitative maintenance procedure,as shown in Fig. 2, with using the proposed
approach.

In this procedure, maintenance engineers
evaluate safety level of the target slope by using the
proposed safety assessment approach, and if it isconsidered that additional surveys are necessary toreevaluate the safety level of the slope, it will beplanned and carried out by maintenance engineers
themselves before consulting to the technicalcommittee. Consequently, it can be expected that theprocedure achieve more rapid and reasonable
maintenance of existing ground-anchored slopes.
The probabilistic safety assessment approachconsists of a probabilistic model to estimate failure
probability of the slope, and its threshold to evaluate
the safety and decide the treatment of the target
slope, which can reproduce the former maintenanceexperience based on experts’ empirical judgments
quantitatively. Fig. 3 shows the image of the model.
In this figure, each plotted point is the combinationof the failure probability of the slope, which isestimated by the proposed probabilistic model, andexcessive tension rate, the ratio between meantensile forces of an excessive tensioned anchor
group and one before rising. If the thresholds drawn
in the figure can reproduce the former maintenanceexperiences based on experts’ empirical judgments,
maintenance engineers will be able to evaluate thesafety level of the target slope and plan the
additional surveys from its failure probability
estimated by the proposed probabilistic model.

Fig. 3 Image of proposal model.
PROBABILISTIC SAFETY APPROACH OF
ANCHORED SLOPE
Slope stability remains safe by ground-anchorsfunctioning against the slope movement. Thisbehavior indicates that tensile forces acting in theanchors of a stable slope continue to decrease from adesign one due to the creep of PC strand and/or
slope ground. On the other hand, the slope stability
with excessive tensioned anchors remains safe byexcessive tensile forces acting in anchors. Namely,
the slope is in the sate in equilibrium of the moving
force and the resistance, the apparent safety factor
should be 1.0. Based on this consideration, safety
evaluations of the slope with excessive tensioned
anchors and one with decreased tension anchors are
assumed as follows;

Fig.2 Future maintenance procedure.

47

GEOMATE- Bangkok, Nov. 14-16, 2016

where, F s is the summation safety factor, F s0 is
the safety factor without considering the effect ofground-anchors, δ FS is the uncertainty of landslide
analysis reproducibility with respect to actual landslide, a uniform distribution that lies between 0.9
and 1.1 [1], T m1 and δ T1 , T m2 and δ T2 are meanvalues and variations of tensile forces acting in eachanchor group, α and θ are the angles between atangential line of each land slide block surface andhorizontal surface, and between anchor installation
angle and horizontal surface (deg), as shown in Fig.
5, respectively, φ is the shear resistance angle (deg),
W is each soil block weight of circular sliding (kN),
T L and δ TL are nominal yield tensile force and its
uncertainty of PC strand of the target anchor definedin quality control document, for instance [2], T F isthe ultimate capacity against pulling out anchor body,which is estimated by T F = U * τ f * L (kN), U is thecircumference of anchor body (m), τ f is the shaft
resistance (kPa), L is the length of anchor body (m),δ FL is the uncertainty of estimation method
reproducibility of T F , m = 1.07, COV = 0.49,
lognormal distribution [3].

Slope with excessive tensioned anchors
Let us consider the case of the slope with two
anchor groups, one consists of the anchors with near
design level tensile forces and another one consistsof the anchors with excessive tensile forces, as
shown in Fig.4. This slope remains safe by theincrease of tensile force, namely, the slope is in thestate in which apparent safety factor is 1.0.
Therefore, the safety level can be evaluated fromfollowing steps.

Fig.4 A slope consisting of two groups of anchors.
Step-1; estimating soil strength parameters, c andφ, from landslide analysis with using mean tension
values of each anchor group, as reproducing thesafety factor of 1.0.
Step-2; estimating failure probability, P f , andactual safety factor of the slope by Eq. (1) withconsidering variations of tensile forces of eachanchor group. Eq. (1) is a performance function,
which means that the summation of the safety
factors of one without considering the anchors’effect and the safety factor gain caused by setting
ground-anchor is larger than 1.0. Furthermore, it wasalso defined that the tensile force acting in theanchors with considering the variation, T m *δ T , does
not exceed yield of PC strand, T L *δ TL , and theultimate capacity against pulling out the anchor body,
T F *δ FL . Here, although the apparent safety factor of
the slope is 1.0 (Pf = 50%) from the viewpoint ofequilibrium, because the probabilities of pulled out
anchor body and broken PC strand exist, larger P fthan 50% and actual safety factor, which is smaller
than 1.0, are obtained. Fig. 6 shows the concept of
the safety evaluation.

Fig. 5 α and θ.

(1)

Fig. 6 Concept of the safety evaluation.

48

GEOMATE- Bangkok, Nov. 14-16, 2016

Slope with decreased tension anchors

VALIDITY OF PROPOSED APPROACH

As mentioned previously, if the slope was instable state, tensile forces acting in anchors would
tend to decrease due to the creep of PC strand and/or
slope ground. Hence the slope with decreased
tension anchors is not in dangerous state. On theother hand, if the decreased tensile forces acting inanchors reascended as shown in Fig. 7, the anchors
would start to function for keeping stability of theslope, namely, the slope is in the state of equilibrium,
the apparent safety factor is 1.0. In this situation, thesafety of the slope has to be evaluated with using thesame way of the slope with excessive tensioned
anchors, in which the tensile forces before
reascending is assumed to be the design one. The
growth of P f , however, is not so large, because thetensile forces acting in anchors is not so high.

The validity of the proposed approach is
confirmed from the application to actual anchoredslopes.
Slope with excessive tensioned anchors
Target slope
The picture of the target slope for confirming thevalidity of the proposed approach is shown in Fig. 8.
In this slope, tensile forces acting in partial anchors
are measured. The measurement of tensile forces iscarried out to 79 anchors (18.7% of 423 entire
anchor numbers). The results and the safety
evaluation area in this confirmation are shown in Fig.
9. The meaning of color in Fig. 9 is summarized inTables 1 and 2. The numbers and the rate of eachrank of measured 79 anchors are presented in Table.
3. The design tensile force of the slope’s anchors is
350 kN.
This slope was judged as a dangerous one andthe installing of additional anchors was proposed asa countermeasure in a technical committee.

Fig. 7 Reascending anchor tension.
Fig. 8 Picture of the target slope.

Fig. 9 Measurement results of tensile forces.
49

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 High tensile force ranking
Rank
A
B
C
D
E

Color
Green
Yellow
Flesh
Red
Brown

Tensile force of anchor
Setting force – Design one– 1.2 times of design force– 1.33 times of design force– 0.9 times of nominal yield
More than nominal yield force

Table 2 Low tensile force ranking
Rank
A
B
C
D

Color
Green
Y-greenBlue
PurpleFig. 10 Difference between apparent and actual ones.

Tensile force of anchor
0.8 – 1.0 times of Setting force– 0.5 times of Setting force– 0.1 times of Setting force
Less than C rank

Table 3 Measured tensile force ranks

High tensile force
Low tensile force

n
%
n
%

A
11
14
11
14

B
13
16
33
41

C
4
5
3
4

D
4
5
-

E
X
X

Safety evaluation
The mean tensile force of upper 9 anchor lines inthe safety evaluation area shown in Fig. 9 is 243 kN
and one of excessive 5 anchor is 424 kN. The soil
strength parameters estimated by landslide analysis
as reproducing safety factor of 1.0 at step-1 are c = 5
kPa and φ = 31 degrees. Then actual safety factor
and failure probability are estimated, as shown in
Table 4 and Fig.10, by Eq. (1) with considering
tensile forces variation of each anchor group shownin Table 5, which are estimated by Monte Carlo
Simulation, MCS, with one million times. Fig. 11
shows the MCS plots to estimate probabilities ofpulled-out anchor body and broken PC strand. From
the results, although the slope remains safe due tothe equilibrium of moving force and excessive
tensile forces, the slope has dangerous of landslide
potentially because of existing probabilities ofpulled out anchor body and broken PC strand.
According to the result, it was considered thatthe proposed approach could reproduce thejudgment made in a technical committee, in which
the slope was judged as dangerous one.

Fig. 11 MCS plots of pulled-out and broken anchor.
Slope with decreased tension anchors
Target slope
Figs. 12 shows the picture of the target slope for
confirming the validity of the proposed approachwith respect to the slope with decreased tension
anchors. Fig.13 shows the tensile force measurementof the anchors composing of the anchored slope andthe safety evaluation area. The tensile force
measurement carried out to 15 anchors (30.6% of 49
entire anchors). The meaning of color in Fig. 13 is
the same as presented in Tables 1 and 2. The design
tensile force of the anchors is 750 kN.
Because the tensile forces of almost anchors are
decreased, C of low tensile force ranking presented
on Table 2, the slope is judged as safe one in atechnical committee and follow-up observation of
the tensile forces acting in anchors has been only
continued.

Table 4 Apparent and actual F s and P f .
Apparent
ActualFs
P f (%)Fs
P f (%)
1.0
50
0.97
66
Table 5 Tension variations of anchors in each group
Group
Mean
Standard deviation
Upper 9 lines
0.69
0.135
Excessive 5 lines
1.22
0.187

Fig. 12 Picture of the target slope.

50

GEOMATE- Bangkok, Nov. 14-16, 2016

out anchor body and broken PC strand, as shown in
Fig. 15 of MCS plots. This result reproduces thepresent judgment made in a technical committee, inwhich this slope is in the state of safety.

Fig. 13 Tensile force measurement results.
Safety evaluation
In order to confirm the proposed approach to thisslope, the occurrence of reascenting tensile force isassumed. Namely, the apparent safety factor of theslope with the tensile forces acting in the anchors
shown in Fig. 13 is assumed as 1.0.
The mean tensile force of the anchors in thesafety evaluation area shown in Fig. 13 is 274 kN.
The soil strength parameters estimated by landslide
analysis as reproducing safety factor of 1.0 at step-1
are c = 13 kPa and φ = 35.5 degrees. Then actual
safety factor and failure probability are estimated, as
shown in Table 6 and Fig.14, by Eq. (1) withconsidering variation of tensile forces acting inanchors, the mean is 0.37 and standard deviation is0.12, which are estimated by Monte Carlo
Simulation, MCS, with one million times.

F.15 MCS plots of pulled-out and broken anchor.
CONCLUSIONS
A probabilistic safety assessment approach forground-anchored slope has been studied in order topropose a future quantitative maintenance procedureof existing anchored slopes. The proposed
probabilistic approach on the present stage of thestudy and the practicability confirmed from theapplication examples to the slope with excessive
tensioned anchors and one with decreased tension
anchors were introduced. As the result, it wasconsidered that the proposed approach could
reproduce the qualitative empirical judgments by
experts. The authors are willing to propose thequantitative maintenance procedure from theapplication of the proposed approach to many
existing anchored slopes and establishment of thethreshold, as shown in Fig. 3, in the near future.

Table 6 Apparent and actual F s and P f .
Apparent
ActualFs
P f (%)Fs
P f (%)
1.0
50
0.999
51

REFERENCES[1] Matsuo, M. and Asaoka, A., “A statistical study
on conventional safety factor method”, Soils
and Foundation, Vol. 16, No.1, pp.75-90, 1976.
[2] VSLA (VSL Association), “A brochure of VSL
permanent anchor method (rank A), 2012.
[3] Okahara, M., Takagi, S., Nakatani, S. andKimura, Y., “A study on bearing capacity of asingle pile and design method of cylinder
shaped foundations”, Technical Memorandumof PWRI, 2919, 1991.

Fig. 14 Difference between apparent and actual ones.
According to the results, the actual safety isalmost same as the apparent one in the case of theslope with decreased tension anchors. This reason isthat the variation of the tensile forces acting in theanchors does not affect the probabilities of pulled-51

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

NEW CONCEPT FOR REASONABLE GEO-DISASTER RECOVERY
Takashi Hara1, Akiho Inoue2, Atsushi Yashima 3, Nobuyuki Soga 4 and Yoshihiro Yokota5
1
Department of Civil Engineering, Gifu University, Japan; 2-4 Gifu University, Japan;
5
MAEDA KOSEN CO., LTD., Japan

ABSTRACT
Various design and construction methods have been proposed and adopted in geo-disaster recovery projects.
However, researches on optimum design and construction methods in geo-disaster recovery projects based on thecomparison of the real recovery projects executed in the respective fields are very few. Because the number ofgeo-disasters, such as land slides, debris flows and so on, is predicted to increase due to the growth of thenumber of sudden heavy rainfalls caused by global climate change, it is considered that the research on optimum
design and construction methods in geo-disaster projects will be necessary from the viewpoint of limited public
investment.
Therefore, the authors have conducted a study on development of a future concept of reasonable geo-disasterrecovery projects based on the real recovery projects executed in the respective fields. The proposed concept andexamples of reasonable disaster recovery project on the present stage of this study as well as each consideration
with respect to geo-disaster recovery in the several fields are discussed in this paper.
Keywords: Geo-Disaster, Recovery, Temporary Structure, Lifelong Duration, Permanent Measure

INTRODUCTION
GEO-DISASTERRECOVERYPRESENTCONSIDERATIONS ON EACH FIELD

The design and construction methods in geodisaster recovery project have been different
depending on the respective fields. For example, inthe field of erosion control, permanent recovery
structure is usually built independently. On the other
hand, the temporal recovery structure is sometimes
applied to a part of permanent recovery structure in
the field of road management. Therefore, thetemporal recovery treatment is called as the
temporal-permanent recovery one in that case.
However, researches studying on reasonable geodisaster recovery projects from the comparison of
the different respective considerations and real
recovery projects executed in the respective fields
are very few. Under the circumstance the increase of
geo-disasters due to the growth of the number ofsudden heavy rainfalls caused by global climatechange, it is considered that the establishment of theconcept of reasonable and optimum recovery geodisaster project is necessary from the viewpoint oftreating the geo-disasters by the limited public
investment.
Therefore, the authors have started conducting a
study on development of a future concept ofreasonable geo-disaster recovery projects from thecomparison of the considerations and real recovery
projects executed in the respective fields. The
proposed concept and examples of reasonabledisaster recovery project on the present stage of thisstudy as well as each consideration with respect togeo-disaster recovery in the several fields are
discussed in this paper.

Erosion Control Field
In the field of erosion control, a temporary
structure for emergent treatment is placed
immediately just after the disaster by using large
soilbags for instance, in order to protect human life
from the occurrence of secondary disaster. Then apermanent recovery structure is planned andconstructed with selecting the appropriate
construction site of the structure, where the national
government or local government purchases thenecessary land for the construction. Namely, theconstruction sites for building temporary structure
and that for permanent structure are in generaldifferent, as shown in Fig. 1.

Fig. 1 Temporary and permanent recovery structure.
Hence, the temporal structure has not been
applied to a part of permanent structure. Therefore,
the temporary structure has been usually cleared52

GEOMATE- Bangkok, Nov. 14-16, 2016

away. The reason is that there is no time to select theappropriate construction site and purchase thenecessary land right after the disaster.

the permanent recovery structure constructed in thepermanent recovery project. In this consideration,
the temporary recovery is called temporarypermanent recovery in the field.
Fig. 2 shows an example of adopting the
temporary-permanent recovery [1-3]. In thisexample, reinforced soil wall composed of large
soilbags and geogrids was adopted as the temporarypermanent recovery structure, as shown in Fig. 2(c).
Then the permanent recovery structure wascompleted by adding another reinforced soil wallcomposed of steel frame and wall strengthenmaterial to the temporary-permanent recovery
structure.

Road Management Field
In the field of road management, building atemporary structure only for temporary recovery
purpose has not been permitted in general.

Afforestation Field
Afforestation structures, as shown in Fig. 3, areplaced to prevent a future erosion and to recover themountain to nature condition. Therefore, thestructures building in the afforestation field arebasically permanent ones and it is not necessary forthe national government or local government topurchase the land for building the structures.

(a) Case of landslide.

(b) Before and after landslide.

Fig. 3 Example of afforestation structure.

NECESSITY OF A NEW CONCEPT FOR
REASONABLE GEO-DISASTER RECOVERY(c) Temporary-permanent recovery.
Variation of Temperature and Rainfall in Japan
According to data of Japan Meteorological
Agency (JMA), the transition of average temperatureof years and the frequency of heavy rainfall morethan 50 mm/hr. observed at 1000 points have been
increasing during recent decades in Japan, as shownin Fig. 4 and 5 [4]. The data indicate that thetemperature raises 1.14 degrees centigrade a hundredyear and the frequency of heavy rainfall more than
50 mm/hr. increases 21.3 times a decade as the trend.
Namely, it is certainly considered that the heavy
rainfall increases with the raise of the temperature in
Japan.

(d) Permanent recovery
Fig. 2 Application of temporary-permanent recovery.
The temporary recovery project is usually planed
to build the temporary structure. At the same time,
its temporary structure is often utilized to a part of53

GEOMATE- Bangkok, Nov. 14-16, 2016

Geo-disaster Scale and Frequency with Rainfall
Figs. 6 and 7 summarize the relationships
between occurrence scale and number of disasterand rainfall per hour with respect to landslide anddebris flow, respectively [5]. The data were obtained
from three prefectures such as Gifu, Toyama andMie, as shown in Fig. 8. Although the data wereobtained from a part of Japan, those can be
representative ones as a mountainous country.
According to the results, both the occurrence
scale and number of debris flows especially increase
sharply with the increase of rainfall per hour. This
result indicates that the occurrence scale and numberof geo-disaster will increase with the increase of thefrequency of heavy rainfall throughout Japan in thenear future.

Fig. 4 Average temperature of years.

Fig. 5 Frequency of larger rainfalls more than
50mm/hr.

Fig. 7 Geo-disaster Occurrence number and rainfall.

(a) Landslide(b) Debris flow
Fig. 6 Relationship between geo-disaster occurrence
scale and rainfall.

Fig. 8 Investigation area.

54

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 9 shows the percentage of emergent
treatments adopted after landslides and debris flows.
Fig. 10 shows the percentage of the failed volume(m3) by landslide with respect to each emergent
treatment method from experiences in Gifu andToyama prefectures [5]. In Figs. 9 and 10, CS iscovering sheet, RS is removing soil, SB is setting
soil bags, Fe is setting fences and Ot is others.

effective to decrease the disaster investment.
Furthermore, the development of the technology for
functionality enhancement and life prolongation of
the structures built as a countermeasure against geodisaster is also important.
A FUTURE CONCEPT OF REASONABLE
GEO-DISASTER RECOVERY PROJECTSThe authors have started a study on thedevelopment of a concept for reasonable geodisaster recovery projects from the viewpoints ofprevious study and preparation, in which thefunctionality enhancement and life prolongation,
development of new recovery methods and theirdesign methods will be discussed. The conceptual
images for road, erosion control dam andembankment studied are shown in Fig. 11.

(a) Landslide(b) Debris flow
Fig. 9 Adopted emergent recovery treatments.

(a) One side filling (b) Both side fillingFig. 11 Conceptual image of temporary andpermanent structures.

As mentioned previously, temporary recovery
structure and permanent recovery one are built on
different place in the erosion control field, and thereason is that there is not enough time to select theappropriate construction site and to purchase theland for building the permanent structure. The
problems, however, can be solved by previous
studies on selecting the construction sites forbuilding permanent recovery structure with theprediction of geo-disaster occurrence prediction andnegotiations with the landowners.
If the prediction of geo-disaster and a-priori
negotiation between officials and landowners are
possible, the application of the temporal-permanent
recovery consideration in the road management fieldto erosion control field will be possible.

Fig. 10 Adopted treatments and scale of debris flow.
Necessity of a New Concept for Reasonable Geodisaster Recovery
Under the circumstance of the increase of theoccurrence number and the scale of geo-disaster in
the near future, the establishment of a new concept
for reasonable and optimum geo-disaster recovery
scheme is very important in order to resist expected
geo-disaster with using limited public investment.
For example, the application of the temporarypermanent recovery consideration in the roadmanagement field to the erosion control field is very

Functionality
Enhancement
andProlongation of Recovery Structures

Life

There have been many cases in which a structure
became to indicate a poor performance after
receiving the geo-disaster impact such as a debris
55

GEOMATE- Bangkok, Nov. 14-16, 2016

flow impact to an erosion dam. Thus thefunctionality enhancement has become one of themost important issues in the erosion control field.
Furthermore, in the afforestation field, it isfundamental that all structures built in the field are
aimed to recover the mountain greenery condition.
The life prolongation of the structures has also
become one of important issues from the viewpointof reasonable afforestation.
Therefore, reasonable methods of functionality
enhancement have been researched in this study. It isvery important to make use of debris flow deposits
for construction material. The authors propose thecomposite structure to enhance the exising dam byusing debris flow deposits, soilbags and concrete
cover, as shown in Fig. 12.

(a) Before landslide(b) Occurred landslide(a) Before debris flow(c) Temporary-permanent recovery(b) After debris flow(d) Permanent recovery
Fig. 13 A simple temporary-permanent recovery.
Designs for recovery methods have been studied.
For example, as far as the design of the permanent
structure by making use of temporary soilbags for anerosion control dam concerned, a design model has
been proposed as the composite element model withelastic beam and compression spring as shown in
Fig.14. A design model shown in Fig. 14 is the caseassumed for verification of internal stability. In thismodel, the covering concrete is modeled as anelastic beam and the interior soilbags are evaluatedas spring with considering long-term compression of
the soilbags. The arrangement of reinforcing bars
should be designed next, as shown in Fig. 15.
Meanwhile, the outer stability is verified from theviewpoints of rotation, sliding and bearing capacity,which is the same as retaining walls. The validity ofa proposed design scheme will be confirmed by(c) Functionality enhancement with debris flow
deposits
Fig. 12 A reasonable functionality enhancement.

Development New Recovery Methods and Their
Design Methods
New reasonable recovery methods, as shown in
Fig.9 for instance, are outlined in this chapter. Fig.
13 shows a simple application of temporarypermanent recovery consideration for a small-scalelandslide disaster, in which large soilbags andcovering soil will be only used.

56

GEOMATE- Bangkok, Nov. 14-16, 2016

Authors first summarized the trend of geodisasters took place in Gifu, Toyama and Mie
prefectures in Japan. The emergent treatments
adopted were alkalized with respect to the disasterscale. Then authors proposed the new concept for
the reasonable geo-disaster recovery in different
fields. The conceptual design images were also

experiments and analyses in the near future.

mentioned for several temporary-permanent
structures. The details of design methods fordifferent structures will be introduced in the near
future.
ACKNOWLEDGEMENTS
The authors would like to acknowledge theengineers of Etsumi-sankei erosion control (SABO)
office of Ministry of Land, Infrastructure,
Transportation (MLIT), Gifu prefecture, Toyama
prefecture, Mie prefecture and Tono Forest Office ofMinistry of Agriculture, Forestry and Fisheries(MAFF), who supplied us valuable data and ideas.

Fig.14 A design model for internal stability.

REFERENCES[1] Kubo, T., Mori, Y. and Miyatake, H., “Thebehavior of the restoration method using thelarge-sized sand-bags”, Geotechnical Journal[2] Mori, Y., Kubo, T. and Miyatake, H., “A
dynamic centrifuge test of disaster recovery
technique with considering site condition”, the70th JSCE annual conference, 2015, pp.699-703.
[3] Kubo, T., Yokota, Y. and Tsuji, S., “Restoration

Fig. 15 An example of rebar’s arrangement.

CONCLUSIONS

and investigation of road embankment damaged
by Noto-Hanto earthquake in 2007”,
Geosynthetics Engineering Journal, 2007, pp.
199-206.

Geo-disaster recovery projects have beenplanned and carried out based on different
considerations depending on several fields, such as
erosion control, road management and so on.
However, because the increase of geo-disasteroccurrence number and growth of the disaster scaledue to the global climate change is estimated, it isnecessary for us to establish a future concept forreasonable geo-disaster recovery project from theviewpoint of efficient use of limited public
investment.

[4] Japan

Meteorological

Agency(JMA),

http://www.jma.go.jp/jma/kishou/books/hakusho/2015/index1.html.
[5] Inoue, A., “A study on efficient and economic
concept of geo-disaster recovery”, Graduation
thesis of Gifu University, 2016.

57

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THE SOIL BLOCK APPROACH FOR VERIFYING HYD USING FEM

1Georgios Katsigiannis1, 2, Pedro Ferreira1 and Raul Fuentes3
University College London, UK; 2 Arup, UK, 3 University of Leeds, UK

ABSTRACTThe HYD limit state, as described in Eurocode 7, is caused by hydraulic gradients and is one of the mostdangerous Limit States, resulting in sudden failure with serious consequences for people and structures. The
hydraulic heave stability problem relates particularly to the upward flow of water through soil to a free surface
such as may occur in front of a retaining wall in the base of an excavation. The simplest way to verify the oil
stability against the HYD Limit State using Finite Element Methods is the so called Soil Block Approach. In thisapproach, which is based on the widely used Terzaghi’s criterion, safety may be checked by studying theequilibrium of a rectangular block of soil. In this paper, the authors describe the Soil Block approach and discuss
its advantages and disadvantages. Comparisons made using a benchmark geometry extensively studied anddiscussed between the members of the EC7 Evolution Group 9 on Water Pressures, illustrate that the calculated
Terzaghi’s factor directly depends on the upstream and downstream groundwater levels as specified by the ratio
Δh/t while for a given difference in the hydraulic head, the system becomes more critical for narrow excavations
where confined space results in an increase in the groundwater pressures. Overall, the authors conclude that the
HYD verification using numerical methods is straightforward and seems very promising.

Keywords: Eurocode 7, HYD, water pressures, FEM

INTRODUCTION
The HYD limit state, as defined in Eurocode 7(EC7), relates to hydraulic heave, internal erosion andpiping in the ground caused by hydraulic gradients[1]. This apparently covers a wide range of situations
with stability problems caused by hydraulic gradients.
This paper focuses on part of this definition, thehydraulic heave stability problem, which is illustrated
in Fig 1. Hydraulic heave refers particularly tovertical upward flow of water to a free surface (e.g.
water flow in front of a retaining wall) and is one of
the most critical Ultimate Limit States, potentiallycausing sudden failures with serious consequences
for both people and structures.
Nowadays, with the increasing popularity of theadvanced numerical methods, more and moredesigners will use Finite Element methods, to verify
soil stability against hydraulic heave. The HYD
verification using FEM can be performed with theSoil Block approach. In this approach, safety isverified by studying the equilibrium of a rectangularsoil mass. [5].
The use of the Soil Block approach has been
recently discussed by the Evolution Group 9, and thiswork draws on those discussions to provide a better
understanding and highlight the feasibility of theapproach for verifying hydraulic heave using
advanced numerical methods.

Fig. 1

Example of a HYD situation.

EUROCODE 7 REQUIREMENTS
Safety against failure by hydraulic heave isverified with Equation 2.9 as given by EC7, where
stability of soil against heave shall be checked either
in terms of forces or stresses [1]. More specifically,
the equation is expressed in two different but
equivalent forms, presented below as Equations 1 and2. The first equation requires the design total verticalstress to be greater than the design pore waterpressure at the bottom of a relevant soil column while

58

GEOMATE- Bangkok, Nov. 14-16, 2016

 Initial stress field conditions are based onhydrostatic water pressures and K0=1-sinφ’.
 Interface elements are used between the soil andthe wall with tanδ = 0.5tandφ’, where δ is the wall
friction angle.

the second requires the design buoyant weight of thecolumn to be greater than the design seepage force.

𝑢𝑑𝑠𝑡;𝑑 ≤ 𝜎𝑠𝑡𝑏;𝑑(1)

𝑆𝑑𝑠𝑡;𝑑 ≤ 𝐺′𝑠𝑡𝑏;𝑑(2)

The soil is homogeneous sand and its
properties are given in Table 2. The generated finite
element mesh is shown in Fig 3.

The subscript d refers to design values ofparameters while the subscripts dst and stb refer todestabilising and stabilising effects.
The partial factor values as specified in UK
National Annex of EC7 are listed in Table 1 [1].

Table 2 Material properties of sandSoil Properties
Young’s Modulus, E' (ΜPa)
25+6.5zAngle of shearing resistance, φ' (°)
35
Effective cohesion, c' (kPa)0
Poisson’s ratio, ν'
0.2*where z is the depth below the ground level (m)Table 1 Partial factors on actions (γF)
ActionSymbolPermanent
Unfavourablea
G;dstFavourableb
G;stb
Variable
Unfavourablea
Q;dsta Destabilisingb Stabilising

Value
1,35
0,90
1,50

C
L

However, it is highlighted in the EC7, that theload factors might not be always suitable for groundwater pressures. The code allows for direct
assessment of the design value or application of asafety margin to the characteristic ground water table.
Thus, by allowing three alternative methods, much of
the responsibility for deriving the design value ofwater pressures is left with the designers [14].
Simpson & Katsigiannis recommend that factoring ofwater pressures should be generally be avoided andsuggest the direct assessment of the design waterpressures or the design water table level [2].

x/2=6mΔh varies

t=3m

HYD VERIFICATION UDING FE METHODS

Fig. 2

Geometry of the cofferdam problem.

Fig. 3Finite Element mesh

Methodology
The Soil Block approach for the HYD
verification using FE methods, is now illustrated forthe simple cofferdam problem presented in Fig 2. The
software used is Plaxis 2015.02 and the following
assumptions were made in the model:
 The wall is wished-in-place, impermeable and notallowed to deform in any direction.
 Only half the excavation width is considered, due
to symmetry.
 Steady state conditions are assumed.
 The side and bottom boundaries are impermeable.
 The side model boundaries are fixed in the x
direction while the bottom model boundary isfixed in both x and y directions.
 The unit weight of the soil γ is equal to 20kN/m3
59

GEOMATE- Bangkok, Nov. 14-16, 2016

The Soil Block Approachfor coarser soils, no explanation of this range has been
given by the authors.
Skempton & Brogan illustrate the significanceof the grading curves of the materials in relation tosafety considerations in presence of hydraulicgradients [9]. Even if water pressures are known withconfidence, the achieved levels of safety highly
depend on the grading curve of the material withpoorly graded materials generally tolerating lower
hydraulic gradients. This is because, in poorly gradedmaterials, the effective stress may vary locally overdistances of the order of a few soil particles, leavingsome particles at much lower stresses than normallycalculated from the depth of overburden.
Similarly, the German guide on erosion makes
a distinction between poorly graded soils that areinternally unstable and well graded soils where thesoil particle mixtures are internally stable [8]. The
critical failure mechanism depends on the grading
curve with internal erosion and particularly suffusion(transport of the fine soil particles through the poresof the coarse particles) being critical for poorly
graded soils and hydraulic heave for well gradedsoils.
This variability of the grading curves and thegoverning failure mechanisms among different soils,
explains why different authors have proposed quite
different values for the Terzaghi’s factor with higher
values typically suggested as an empirical way to
account for the composition anomalies of internallyunstable soils.
It is generally recommended that designers
first investigate if the soil is internally stable before averification of hydraulic heave can be done rather
than generally use increased safety factors for theverification of hydraulic heave.

The Terzaghi’s criterion
Terzaghi was the first to propose study of a soil
block extending a depth t from the free surface to thetoe of the wall and of width b=t/2 for isotropic anduniform materials [3], [4]. According to Terzaghi’s
criterion, soil stability is checked by verifying that thebuoyant weight of the block is greater than theseepage force (see Fig. 4). The friction on the sides ofthe block at its interfaces with the wall and with therest of the soil is ignored. Terzaghi’s factor of safety
is defined in Equation 3 below, where W is the weightof the block, H is the force on base of block due tohydrostatic pressure, U is the water force on base ofblock, W-H is the buoyant weight and U-H is theseepage forces.

𝐹𝑇 =

𝑊−𝐻
𝑈−𝐻

(3)

Although Terzaghi et al. give a worked
example in which the acceptable factor required is
FT=2.5 [10], no direct recommendation from
Terzaghi has been found, in previous publications,with the specification of a minimum factor of safety.
Values taken from a survey of publications, generally
based on the use of Terzaghi’s diagram, aresummarised in Table 2 [2].
Table 2. Published values for Terzaghi’s factor ofsafety FT [2]Publication and any limitations

ValuesWilliams & Waite (1993)

1.5 to 2.0

The Soil Block Approach with FEM
Calculating the Terzaghi’s factor with FE
methods is straightforward as the weight of the soil
block and the hydrostatic pressure on the base of theblock can be easily derived from hand simple
calculations. The pore water pressures acting at thebottom of the block are obtained from the output of
the FE calculations.
As mentioned before, Terzaghi recommended
that a column of width b=t/2 should be used in thecalculations of the factor of safety, taking no accountof friction forces on its vertical sides. Simpson &Katsigiannis showed, however, that the hydraulicgradient would be higher if a narrower column wereused [2]. It could be that Terzaghi considered that anarrower column is unlikely to fail because thefavourable effect of the friction forces on its verticalsides would become significant. For this study, all thesoil block calculations are based on the Terzaghi’s
block dimensions; the depth of the block is equal tothe embedment depth t and the width b is equal to t/2.

For clean sands
Kashef, Abdel-Aziz Ismail (1986)

4 to 5Harr (1962)– unfavourable soils

4 to 5
1.9(DIN 1054/A2 2014) – favourable soils

1.42

Swedish practice – coarse soils

1.5(Ryner et al 1996) – silty material

2.5

Dutch practice2.8

Das (1983), quoting Harr (1962)

4 to 5German practice

The values for the required factor of safety
shown in Table 2, range from 1.42 to 5. While someauthorities require larger factors for finer soils than60

GEOMATE- Bangkok, Nov. 14-16, 2016

As the buoyant weight, which is the stabilisingforce, only depends on the unit weight of the soil, γ
and can be easily calculated for the Terzaghi’s blockas defined in Figure 3.5, the Terzaghi’s factor is moresensitive to variations of the destabilising force whichis the seepage force caused by the pore waterpressures. The effects of different parameters on thepore water pressures and hence the Terzaghi’s factor,
are investigated in this study.

H1

b=t/2

H2

G’

t

Permeable layer
Possiblefailurezone

Fig. 4S

Fig. 5

Calculated Terzaghi’s factor for varying Δh/t

Fig. 6

Head equipotentials for a) Δh/t=1, b)Δh/t=1.5 and c) Δh/t=2

The Terzaghi’s criterion.

Effect of Δh/t
The calculation of the factor is now illustrated
for a simple cofferdam problem for varying Δh/t ratio,
where Δh is the difference in the hydraulic head and t
is the embedment depth of the wall. By gradually
increasing the Δh/t ratio, the analysis was driven to
failure. Different hydraulic heads were achieved by
specifying different levels of the free-water tablebehind the retaining wall and at the end of eachanalysis, the Terzaghi’s factor was calculated. In
Figure 5, the factor is plotted against the ratio of Δh/t
and it can be seen that its value drops as Δh/t
increases. This is because an increase in Δh gives riseto the groundwater pressures as shown in Fig 6. The
system becomes unstable and the value of the factor
less than unity for a ratio of Δh/t equal to 2.25. The
corresponding vertical soil displacements vectors arepresented in Fig. 7 where the well-defined failuremode can be observed. The results are consistent withthe failure mechanisms presented by other authors for
similar problems [12], [13].
It can be concluded that the calculated
Terzaghi’s factor directly relates to the specified
upstream and downstream groundwater levels as
expressed by the ratio Δh/t.

Fig. 7 Vertical displacement vectors for Δh/t=2.25
Effect of x/t
61

GEOMATE- Bangkok, Nov. 14-16, 2016

Simpson & Katsigiannis also illustrated theeffect of the excavation width when calculating the
Terzaghi’s factor [2]. When plotting the headequipotentials for wide and narrow excavations, theauthors observed that pore water pressuressignificantly rise when for small excavation widths.
To better illustrate this effect, the analysis isrepeated for different excavation widths while the restof the model parameters remain the same. Morespecifically, 5 different cases were considered forplain strain conditions: x/t=12, 8, 4, 2 and 1, where x
is the excavation width (due to symmetry only half
the excavation is modelled) and t is the embedment
depth. At the end of each analysis, the Terzaghi’s
factor was calculated using the values of the porewater pressures acting at the bottom of the soil blockfrom the output of the calculations. This study
includes 5 different geometries each simulated using
three different values of Δh/t, totalling 15 analyses.
In Figure 8, the Terzaghi’s factor is plottedagainst the ratio x/t for Δh/t=1.5. It can be seen, thatthe narrower the excavation is, the lower the factor ofsafety becomes. The value of the factor of safety
particularly drops for values of x/t lower than 4.
Figure 9 gives the head equipotentials for varying thex/t parameter. It can be seen that the ground waterpressures rise significantly as the excavation becomesnarrower. This rise in the water pressures causes thedrop in Terzaghi’s factor value. This confirms thefindings of Aulbach & Ziegler that when water isflowing upwards beneath a narrow excavation, theupward hydraulic gradients are higher than in thecases of wider excavations with little or no lateral
restraint [11].

Fig. 9Head equipotentials for x/t a) x/t =1, b) x/t =4
and c) x/t =12

CONCLUSION
Verification of stability against HYD using FE
methods is straightforward and seems very
promising. The Soil Block approach is simple andrequires only the pore water pressures acting at thebottom of the soil block to be calculated from the FE
analysis.
The calculated Terzaghi’s factor directly
depends on the upstream and downstream
groundwater levels as specified by the ratio Δh/t. It
was also noted that for a given difference in thehydraulic head, the system becomes more critical for
narrow excavations where confined space results in
an increase in the groundwater pressures.
While the Soil Block approach has the benefit ofsimplicity, it has the disadvantage that it can beapplied only to the simple problems similar to the one
studied in this paper, and not to more complex
geometries such as slopes and embankments.
Moreover, it provides no useful information about thesoil stability at a very small scale but instead it predefines the critical failure mode.
In general, for HYD verifications it isconsidered good practice not to merely rely on factorsof safety. Designers should always be aware that evensmall variations in permeability, which could very
easily be overlooked in a ground investigation, could
lead to instability [2].
ACKNOWLEDGEMENTS
The authors gratefully acknowledge the supportof the project partners, EPSRC, Arup and UCL
European Institute. Special thanks are extended to
Brian Simpson from Arup Geotechnics for the useful
discussions.
REFERENCES[1] BS EN 1997-1 (2004). Eurocode 7 –Geotechnical design, Part 1 – general rules,
British Standards Institution, London, 999pp

Fig. 8Calculated Terzaghi’s factor for varying x/t62

GEOMATE- Bangkok, Nov. 14-16, 2016[9] Skempton, A. W., & Brogan, J. M. (1994).
Experiments on piping in sandy gravels.
Géotechnique. 44, No. 3,449-460.
[10] Terzaghi, K., Peck, R., & Mesri, G. (1996). Soil
Mechanics in Engineering Practice, 3rd edition.
John Wiley & Sons, Inc.
[11] Aulbach, B., & Ziegler, M. (2013). Simplified
design of excavation support and shafts for safety
against hydraulic heave. Geotechnics andTunnelling, 362-374.
[12] Benmebarek, N., Benmebarek, S., & Kastner, R.
(2005). Numerical studies of seepage failure ofsand within a cofferdam. Computers andGeotechnics, 264-273.
[13] Stelzer, O., & Odenwald, B. (2015). A New
Stress-Based Approach for Verification of Safety
Against Hydraulic Heave Based on EC7. XVIEuropean Conference on Soil Mechanics andGeotechnical Engineering (pp. 4073-4078).
Edinburgh: ICE Publishing.
[14] Simpson, B., Vogt, N. and Van Seters, A.J.
(2011). Geotechnical Safety in Relation to Water
Pressures. Proceedings of the 3rd International
Symposium on Geotechnical Safety and Risk.
Munich.

[2] Simpson, B. and Katsigiannis, G. (2015). Safety
considerations for the HYD limit state. XVIECSMGE, Edinburgh 2015[3] Terzaghi, K. (1922). Der Grundbruch an
Stauwerken and seine Verhiltung, Die
Wasserkraft, Vol. 17, No. 24, pp. 445-449[4] Terzaghi, K. (1943). Theoretical Soil Mechanics,
J. Wiley and Sons, New York[5] Evolution Group 9 - Water Pressures. (2014).
Final report to TC250/SC7 Evolution ofEurocode 7 Part 1. July 2014[6] Stelzer, O., & Odenwald, B. (2015). A New
Stress-Based Approach for Verification of Safety
Against Hydraulic Heave Based on EC7. XVIEuropean Conference on Soil Mechanics andGeotechnical Engineering (pp. 4073-4078).
Edinburgh: ICE Publishing.
[7] Simpson, B. (2012). Eurocode 7 – fundamental
issues and some implications for users. 16th
Nordic Geotechnical Meeting (pp. 29-52).
Copenhagen: Danish Geotechnical Society.
[8] BAW (2013). Code of Practice: Internal Erosion(MMB).
GermanFederal
Waterways
Engineering and Research Institute, Karlsruhe

63Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

IMPROVEMENT OF HIGH EXPANSIVE SOILS BY DEEP SOIL
MIXING METHOD IN THE SMALL SCALE LABORATORY
EXPERIMENT
Yulvi Zaika1 and Arief Rachmansyah2
Faculty of Engineering, Brawijaya University, Indonesia

1,2

ABSTRACTThe expansive soil could damage on the pavement and other floor building because of its swellingcharacteristic, low strength and California Bearing Ratio. Improvements by adding stabilizing material is one
alternative that could reduction swell potential and increasing of CBR. Because the thickness of expansive soils
in Bojonegoro District, East Java, Indonesia are more than 1.5 m , the improvement by adding stabilizing
material in the surface is less effective. Deep Soil Mixing (DSM) is ground modification technique can be done
to depth up to 30 m. This paper will explain the result of laboratory experiment of deep soil mixing (DSM)
method as suitable alternative. The small-scale experiments will be applied to observe the change of strengthand swell potential. Implemented DSM requires consideration of the stabilized material (binder) with optimum
value, water – binder ratio that provides maximum performance on DSM and geometric parameters (length,
diameter and spacing of columns). The fly ash material will be used as the material will be made a series ofexperiments to get the optimum value. The optimum water content of expansive soil will be used and theconfiguration single square were implemented. The experiment showed that the strength of expansive soil
increase and the swell potential decrease. The effect of DSM could increase the soil strength and reduce its swellpotential.
Keyword: expansive soil, Deep Soil Mixing, swelling, strength

INTRODUCTION

river
Bengawan
Solo and sedimentation in wetlands (Qa). Thissoil is generally fine-grained, and contains a lotof organic matter. The thickness of young soft
sediment soil can reach 40 m depth (Wesley, 2010).
The second type of clayey soil is the product ofweathering process of mudstone (residual soil) of
the Lidah Formation. The Lidah Formation (TQs)
composed by blue grey mudstone with lenses ofcarbonatics sandstones and limestone. Thisgeological formation is formed in a shallow marine
environment during the Pliocene - Pleistocene(Ratman, et al., 1998). This residual soil type has
generally high shrink-swell characteristics (known
as high expansive soil), because of its high contents
of montmorilonite minerals. Bojonegoro district is
one of area in this formation. The expansive
characteristics of the clayey soil can cause cracks
in the floor of building and the road, and prone thelandslide (Fig. 2).

One of geotechnical problems encountered in
the construction of transportation infrastructure in
the north part of East Java is wide spread expansive
clayey soil (Fig.1). In this region it is founded
two types of clayey soil.

Fig.1 Map of the spread of soft soil in the northernpart of East Java, Indonesia and transportation
infrastructure (Rachmansyah, 2014).
The first one is the sediment soil that as the
product of sedimentation materials carried by the64GEOMATE- Bangkok, Nov. 14-16, 2016
Fig. 2 Typical cracks of the pavement that werebuilt on the residual soil of Lidah Formation in the
North East JavaTo improve the soil properties, in this case to
reduce shrink swell characteristics are usedcommonly cement, lime, silica and various
chemicals compounds. In an industrial context, fly
ash usually refers to ash produced during
combustion of coal. Fly ash is generally captured
by electrostatic precipitators or other particle
filtration equipment before the flue gases reach thechimneys of coal-fired power plants. Because of its
contains such as silica (SiO 2 ), iron oxide (Fe 2 O3),
aluminium oxide (Al 2 O 3 ), and calcium oxide(CaO), fly ash can be categorized as pozzolanicmaterials.
Considering the successful and rapid nature ofchemical based DSM in stabilizing soft soil and thesame technology was considered to expansive soil.
The purposed of this study was to explore thechange of strength and swell potential of
improvement with DSM in difference geometricparameters (length, diameter and spacing ofcolumns) in small scale laboratory experiment.
Laboratory result on field cores indicated that bothfield stiffness and strength are about 20% to 40%
less than the corresponding laboratory prepared soil
samples (Madhyannapu et al, 2010). Some ofnumerical studies have been carried out tounderstand the behavior column-supported
embankment system, assuming linear elastic orlinear elastic perfectly plastic but Yapage et al.
(2014) obtained
strain softening constitutive
behavior for cement stabilized clay for DSM
column. Application and evaluation of DSM wasdeveloped by Madhyannapu at al.(2014) to propose
the stepwise design methodology.

the soil sample compacted at optimum moisture
content in a consolidation ring and soaked under asurcharge of 6.9 kPa (1 psi) , Seed et al. (1962).

𝑆𝑆𝑆𝑆 =

𝑞𝑞𝑢𝑢

𝑥𝑥100%

Deep Soil Mixing (DSM)
The use of the soil mixing to improve theengineering properties of soft soil andcontaminated soil is very well known for a longtime. In this method soils are mixed in situ withstabilized binders. The stabilized material willproduced the higher strength, lower permeability,
and lower compressibility and lower swell potential
than native soil. The improvement becomes
possible by cation exchange at the surface clay
minerals, bounding of soil particles and/or filling ofvoid by chemical reaction product. The widelyused binder are cement and lime, but slag, gypsum,
fly ash and secondary products are also used.
The soil mixing can be divided into two general
methods: the deep soil mixing and shallow soil
mixing. The DSM is an applied for in situ
stabilization of soil to minimum depth of 1.5 m(depend on mass stabilization). Generally DSM
can be performed in two techniques of mixing, dry
mixing and wet mixing. In the wet mixing, the soil
is mixed with additives that have been in slurry,
while the dry mixing soil mixed with additives indry conditions.
The model for predicting the heave of
expansive subsoil was based on the variation ofswell pressures with depth and is presented in thefollowing equation (Fredlund and Rahardjo 1993):

How DSM improve soil characteristic is wellunderstood by using Bearing Capacity Ratio or
Bearing Capacity Improvement (BCI) that can be
determined based on ultimate bearing capacity andbearing capacity of treated soil (DSM).
Formulation of BCI can be written as
𝐵𝐵𝐵𝐵𝐵𝐵 =𝐻𝐻𝑖𝑖(2)
where, H i : initial height of sample, H f : final
heightSwell potential was percentage of swell under1-psi surcharge of sample compacted at optimum
water content to maximum density in Standard
AASTHO Compaction Test. Set the swell gauge tozero and record the time of the start of test. The test
will be stopped when the swell terminated.

The Bearing Capacity Improvement

𝑞𝑞𝑢𝑢(𝐼𝐼)

𝐻𝐻𝑖𝑖 −𝐻𝐻𝑓𝑓∆ℎ = ∑𝑛𝑛𝑖𝑖=1

(1)

(3)

Where:qu (I) : ultimate bearing capacity with DSM
qu : ultimate bearing capacity native soil

𝐶𝐶𝑠𝑠,𝑖𝑖 ℎ𝑖𝑖

1+𝑒𝑒0,𝑖𝑖

log

𝑝𝑝′𝑓𝑓,𝑖𝑖
𝑝𝑝′𝑠𝑠,𝑖𝑖

Where:
C s,i : swell index
e o,i : initial void ratiop’ f,i : final stress (overburden ± any change of total
stress)p’ s,i : initial swell pressure (suction).
h i : thickness of each layer

Swell Potential
Determination of swell potential is quite
important in design of foundations on expansive
soils. The swelling tendencies of expansive soils
are quantified by the swelling potential and/or freeswell. Swell potential or volume change is definedas the ratio of increase in high to the initial high of

According to Rao at. al. (1988) in unsaturated
expansive soil the initial swell pressure could bemeasured as the corrected swell pressure, p’ s,i from65

GEOMATE- Bangkok, Nov. 14-16, 2016
Volume Ratio (stabilized soil ratio)

the constant volume type oedometer test. The final
stress state, p’ f,i , accounts for the overburden stress,
as well as any net change in total stress from either
excavation or surcharge type loading. The Eq. (3)was modified into Eq. (4).
𝐶𝐶

ℎ

𝑖𝑖
∆ℎ = ∑𝑛𝑛𝑖𝑖=1 𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖
𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐 log

1+𝑒𝑒0.𝑖𝑖(4)

Estimate the stabilized area or volume ratio (ifthe treatment was not end bearing column) is
required to reduce overall swelling to tolerable
swelling (heave).
The diameter of DSM column depend on theavailability of local DSM rigs. The area ratio ofsquare arrangement could be calculate based on Eq.
10.

𝑝𝑝′𝑓𝑓,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖
𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖

𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐, 𝑝𝑝′𝑓𝑓,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖
Where the parameter 𝐶𝐶𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖 , 𝑒𝑒0.𝑖𝑖
and 𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖 are composite properties of layer i in
the treated ground. The parameter could be
estimated as shown following:𝑎𝑎𝑟𝑟 =(10)

𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,1 = 𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐 . 𝑎𝑎𝑟𝑟 + 𝑝𝑝′𝑠𝑠,𝑠𝑠𝑠𝑠𝑠𝑠𝑠𝑠 . (1 − 𝑎𝑎𝑟𝑟 )
(6)

(7)

=1

1+𝑒𝑒0.

log=

𝑠𝑠𝑐𝑐𝑐𝑐 .𝑠𝑠𝑐𝑐𝑐𝑐

4

MATERIAL AND METHOD
Soil Properties

Where subscript soil= untreated soil properties andsubscript col = treated soil.
Because of initial void ratio and bulk unit
weight for treated and untreated are the same andconstant with depth and the composite properties
𝐶𝐶𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖 , 𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖 were also constant with depth,
the Eq.(4) became more simple:
𝐶𝐶𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖 ℎ𝑖𝑖

𝑎𝑎𝑠𝑠𝑠𝑠𝑠𝑠𝑠𝑠 +𝑎𝑎𝑐𝑐𝑐𝑐𝑐𝑐

2𝜋𝜋𝑑𝑑𝑐𝑐𝑐𝑐
𝑐𝑐�

where a col = area of column DSM, d col = diameterof column DSM and s cc is space between 2columns. Volume Ratio is calculated by area andlength of column.

𝐶𝐶𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,1 = 𝐶𝐶𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐 . 𝑎𝑎𝑟𝑟 + 𝐶𝐶𝑠𝑠,𝑠𝑠𝑠𝑠𝑠𝑠𝑠𝑠 . (1 − 𝑎𝑎𝑟𝑟 )(5)∆ℎ = ∑ 𝑛𝑛𝑖𝑖

𝑎𝑎𝑐𝑐𝑐𝑐𝑐𝑐

The characteristics of expansive soil in this study
were shown table 1.
Table 1. Characteristics of expansive soil
Characteristic
Value
Liquid Limit (LL)
73.9%
Plastic Limit(PL)30.4%
Shrinkage Limit(SL)2.8%
Plasticity Index (PI)
43.1%
USCS
CH

𝑝𝑝′𝑓𝑓,,𝑖𝑖

𝑝𝑝′𝑠𝑠,𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐,𝑖𝑖

Based on specification of PI and LL wasproposed by Horz (1956), Raman (1967), Chen
(1988) and Snethen (1977) the degree expansion ofsoil is high.

Bearing Capacity of Composite Ground
Besides the compressibility, the improvement
bearing capacity of expansive soil also important inapplication DSM. Some methods to calculate
bearing capacity of composite ground (end bearing
column) are expressed:(1) Weighted method

Box Model and Soil Preparation of Loading Test
Prime element that used is box, made of fiber
glass with length 100 cm, width 50 cm and height
30 cm. The box use L profile as the frame and it
made to be rigid enough for maintain strain planecondition. Fiberglass used as box to makeobservation easier in laboratory (Fig.3).
The loading test performed on the untreated soil
with 27.9 % moisture content and dry unit weight
1.288 gr/cm3 and treated soil in 25.8% watercontent and 15% fly ash (based on previous
research). Compaction is controlled from weight ofsoil that is loaded into the box with a volume that isrequired. The fly ash and soil are mix in dry
condition.
The holes in treated soil test are made by using
a steel pipe of the same size with a diameter of
DSM column (2cm). In this experiment attempted
to vary spacing and depth to the composition of asingle square (Fig.4)

𝑞𝑞𝑢𝑢 = 𝑐𝑐𝑢𝑢𝑢𝑢 𝛼𝛼 + (1 − 𝛼𝛼)𝑐𝑐𝑢𝑢𝑢𝑢(8)(2) Broms (2000); Bouassida and Porbaha (2004)
Method𝑞𝑞𝑢𝑢 = 0.7𝑞𝑞𝑢𝑢𝑢𝑢 𝛼𝛼 + 𝜆𝜆(1 − 𝛼𝛼)𝑐𝑐𝑢𝑢𝑢𝑢(9)Where:
c uc : undrained shear strength of column
c us : undrained shear strength of soil(untreated)
q uc :unconfined compressive strength of column
α: repalcement area ratio
λ : 5.5 (Bergado at al., 1994)

66

GEOMATE- Bangkok, Nov. 14-16, 2016

cohesive (cu) after improvement
sensitivity of expansive soil.

and

also

of

soil

RESULTS AND DISCUSION
Cohesive
Un-drained
Improvement

Strength

Clay soil exhibit significant increase in strength
when treated with fly ash. The table 3 shows themeasured value of un-drained shear strength inundisturbed, remolded, optimum water content
without treated and treated sample.
Fig. 3 Box experiment and loading test.

Table 3 Un-drained shear strength of suchcondition expansive soil
SampleWater content cu (kg/cm2)(%)
undisturbed
49.46
0.253
Compacted soil 27.8 (OMC)
0.643
Treated soil
25.8
0.981

Treated soil is inserted into the hole withseveral layers. Each layer was compacted. The
number of collisions is done by way of apreliminary study to obtain desirable density andcuring in 3 days.

The strength of expansive soil increase 155% bycompaction method and increase 289% by mixed
by fly ash in optimum content.
Column DSM

Spacing and Length Column Effect to Strengthand Settlement

Plate

One the most important analyses of DSM iscontrol of settlement of structure. Fig. 5 shown therelation between stress and settlement of soil. The
untreated soil reached ultimate stress at 380 kN/m2.
At the same stress, the treated soil by DSM column
establish small settlement. It made sense the treatedsoil was stiffer than untreated one.

Fig. 4 Single square model
Variation of length and spacing of columns can beseen in Table 2.
Table 2 Length and Spacing of columns
Parameterscm
Length (L)B, 2B, 3B
Spacing(S)D,1.25D, 1.5D
Where: B= width of loading plate (square) = 5 cm,
D= diameter of column=2 cm.

Settlement (mm)0Soil Preparation and Swelling Test
Swelling tests were applied for untreated andtreated soil in the same condition (water content,
dry density, cure time) in CBR mold. There are 1, 2,
3, 4, 5 and 6 column are made by treated soil andapplied swelling test. The test is conducted toobtain relationship between volume ratio of
stabilized soil and swelling potential. The relation
will be used to estimate swell potential in small
scale model.

0
0.5
1
1.5
2
2.5
3
3.5
4
4.5
5
5.5
6
6.5
7
7.5
8
8.5
9

100

200

300

400

stress (kN/m2)
500 600

700

800

900

1000 1100

untreated soil
S=D; L=B
S=1.25D; L=B
S=1.5D: L=B
S=D; L=2B
S=1.25D;L=2B
S=1.5D; L=2B
S=D; L=3B
S=1.25D; L=3B
S=1.5D; L=3B

Fig.5 Relation between stress and settlement oflength and column spacing of DSM
The longest and closest column reached peakload value. The result showed linear elastic
perfectly plastic behavior of DSM columns. Basedon Euro Soil Stab (2012), the curve is linear up to
the long term strength (creep strength) and then theexceeded load is assumed to be constant as shownin Fig.6.

Unconfined Compressive Strength Test
Unconfined Compressive Strength tests wasapplied to investigate the change of un-drained67

GEOMATE- Bangkok, Nov. 14-16, 2016

1200

qu(kg/cm2)

1000

σcol
Ecol

800
600
y = 0.0115x2 + 9.7956x + 404.26

400
200

σcreep (creep strength)0
0.0

Deformation,ε

20.0

40.0

60.0

80.0

Ratio of stabilized soil (%)

Fig. 6 Stress and deformation in column stabilized
soil (Euro Soil Stab)Fig. 8 Influence Stabilized Area to Bearing
Capacity

Bearing Capacity Improvement (BCI)The Swelling Potential of Improvement Soil

Bearing capacity improvement is a measure
how high the improvement method to increase
ultimate bearing capacity. Based on 3 variations of
spacing and length of column, relation between
Bearing Capacity Improvement and the parameters
could be analyzed. The experiment give the result
the highest BCI is 2.73 for the longest and nearest
column spacing and the lowest one is 1.147 for the
shortest and widest space column (Fig.7).

The swelling potential is comparison of volume
change to initial volume in percentage. The
observation of swell potential for DSM method wasdone by variation of number of column soil
improvement. In this case the volume of soil
improvement was change and the swell potential
are measured.
The relationship between ratio of soil
improvement volume and swell potential shown in
Fig. 9.

3.0

4.50

2.6

L=B

2.4

L=2B

2.2

L=3B

Swell Potential

BCI

2.8

2.0
1.8
1.6
1.4
1.2

4.00
3.50
3.00
2.50

y = -0.0011x2 - 0.023x + 4.0799
R² = 0.9763

2.00
0 2 4 6 8 10 12 14 16 18 20 22 24
Ratio of stabilized soil (%)

1.0
0.75

1

1.25

1.5

1.75

S/D

Fig. 7 BCI of soil for difference space and length of
DSM column
Fig. 8 Ratio of improvement soil and swellpotential

Length and spacing of the columns have highly
impact affect to the area or volume of soil
improvement so as to give effect to an increase in
the strength capacity of the soil as show in Fig.8. In
this study the relation between volume of stabilized
soil and bearing capacity was nonlinear. But basedon Eq. 8 and 9, relations between bearing capacity
and the ratio of the stabilized soil is linear for end
bearing column.

The regression analysis was applied to data sample
and gave the formulation as follows:
𝑦𝑦 = −0.0011𝑥𝑥 2 − 0.023𝑥𝑥 + 4.0799(11)

The equation was used to observe swellpotential of DSM in scale model. The increasing
ratio of volume of stabilized soil will decreaseswell potential. There are some factors influence todecrease of swell potential such as characteristic
expansive soil, type of binder, water content,
curing time and mixing methodology.

68

GEOMATE- Bangkok, Nov. 14-16, 2016

The estimation of swell potential of DSM atsquare arrangement could be read in Table 4.

[2] Bergado, D.T, Chai, J.C, Alfaro, M.C andBalasubramaniam,
A.S,”
Improvement
techniques of soft ground in subsiding andlowland environment”, 1994, Balkema,
Retterdam, 1-220.
[3] Boussida, M, Porbaha,”Ultimate bearing
capacity of soft clays reinforces by a group ofcolumn application to a deep mixing
technique”, Soils and foundations, 2004,
vol.44, No.3, pp. 91-101.
[4] Broms, B.B,” Lime and lime/columns,
summary and visions, Proceeding 4th
international
conferenceongroundimprovement geo-system, 2000, vol. 1, 43.
[5] Chen, F.H.” The use of piers to prevent the
uplifting of lightly loaded structure foundedon expansive soil, Concluding Proc. Eng.
Effects of moisture change in soils, Int. Res.
Eng. Conf. expansive clay soil, 1965, Texas
A&M Press, pp.152-171.
[6] Euro Soil Stab,” Development of design andconstruction methods to stabilize soft organic
soil: Design guide soft soil stabilization,
CT97-0351, Project no.BE-3177[7] Fredlund, D G, Raharjo, H,” Overview of
unsaturated soil behavior”, Proc. Unsaturated
soil, 1993, ASCE, pp. 1-31.
[8] Holtz, W.G and Gibbs, ”Engineering
properties of expansive clays’, Transaction
ASCE, 1956, Vol.121, Issue 1,pp. 641-677.
[9] Madhyannapu, RS and Puppala, AJ, Nazarian,S, Yuan, D ,”Quality assessment and quality
control of deep soil mixing construction forstabilizing expansive sub soils ”, J.
Geotechnic & Geo-environment Engineering,
10.1061/(ASCE)
GT.1943-5606.0000188,
2010, pp. 119 – 128.
[10] Madhyannapu, RS and Puppala, AJ ,”Designand construction guidelines for deep soil
mixing to stabilize expansive soil”, J. Geotech.
Geoenviron. Eng., 10.1061/(ASCE)GT.19435606.0001149, 2014, Pp. 04014051-104014051-14.
[11] Rahmansyah,A,” Improvement of expansive
soil of east java, Indonesia by using industrial
solid waste”, Proc. Int. Conf. on unsaturated
soil UNSAT, 2014, Sydney, Australia.
[12] Raman,V,” Identification of expansive soils
from plasticity index and the shrinkage index
data”, Indian Eng., 1967, Calcuta 11(1), pp.
17-22.
[13] Ratman, N, Suwarti, T, and Samodra, H,“Peta Geologi Indonesia, Lembar Surabaya,
skala 1:1000.000, Pusat Penelitian dan
Pengembangan Geologi, 1998, Bandung(indonesia version)[14] Rao, R.R, Rahardjo,H, Fredlund, D.G, Closed
form heave solutions for expansive soils,

Table 4 Swell potential of square arrangement
DSM
Spacing
Length ofof column
columnuntreated soil

Ratio of
Stabilized soil
0.00

Eq.11
4.08

D
2D
3D
D
2D
3D

B
B
B
2B
2B
2B

19.63
12.57
8.73
39.27
25.13
17.45

3.20
3.62
3.80
1.48
2.81
3.34

D
2D
3D

3B
3B
3B

58.90
37.70
26.18

-1.09
1.65
2.72

The tolerable heave for flexible pavement isgenerally on the order of 12 mm and the swellpotential for untreated soil is 4.08%. If it assumed
the depth of potential swelling in field is 1.5m=
150 cm and the ratio of stabilized was 45% then theswell potential was 0.8%, the tolerable condition
can be reach.

SUMMARY AND CONCLUSION
This study is research a small scale inlaboratory to evaluate DSM construction forstabilizing expansive soil to predict the condition in
actual field.
Prior to DSM construction, laboratory mix designneed to conduct such as swelling and strength
untreated and treated soil in optimum dosage ofbinder.
The variation of space and depth of column are
directly influence to ratio of stabilized soil and also
strength and swell potential. The relation is not in
linear function in which increasing of ratio will
increase strength and decrease swelling.
To choose geometric parameter such as
diameter and depth in installation pattern (square,
triangular), it is needed study intensively aboutswell potential, binder content and also watercontent in field.

REFERENCES[1] Asano, J, Ban, K and Azuma, K,” Deep
mixing method of soil stabilization using coalash”, Proc. The second international
conferenceongrouting
improvement
geosystem, Balkema Rotterdam, 1996,
pp.392-398.

69

GEOMATE- Bangkok, Nov. 14-16, 2016

Journal Geotecnical Engineering, 1988,
Vol.114, No.5, Pp. 573-588.
[15] Snethen, D.R, Johnson,L.D and Patrick,D.M,”
An evaluation of expedient methodology for
identification of potentially expansive soil,
soil and pavement lab.”, US Army Eng.,1977,
Waterway Exp.Sta, Vicksburg, MS,Rep. No.
FHWA-RE-77-94,NTIS, PB-289-164.
[16] Seed,H.B, Woodward. R.J and Lundgren, R,”
Prediction of swelling pontential forcompected clay”, Journal ASCE, Soil
Mechanics
andFoundation
Division,
1962,Vol. 88, Issue 3, pp. 53-88.

[17] Wesley, L D, “ Fundamental of Soil
Mechanics for Sedimentary and Residual
Soils, 2010,John Wiley and Sons
[18] Yapage, N.N.S, Liyanapathirana, DS, Kelly,
RB, Poulos, HG, Leo, CJ,” Numerical
Modeling of an Embankment over soft ground
improved with deep cement mixed column:case story”, Journal Geotech. Geoenviron Eng.
10.1061/(ASCE)
GT.1943-5606.0001165,
2014, pp. 04014062-1-04014062-10.

70Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

3D-FE ANALYSES OF PILED RAFT FOUNDATION WITHCONSOLIDATION IN BANGKOK SUBSOIL CONDITION
1,3

K. Watcharasawe1, P. Kitiyodom2 and P. Jongpradist1
Faculty of Engineering, King Mongkut's University of Technology Thonburi, Thailand
2
Geotechnical & Foundation Engineering Co., Ltd. (GFE), Thailand

ABSTRACTThis study focused on the investigation of influencing factors on behaviour of piled raft foundation in Bangkok
subsoil. To evaluate the possibility of implementing this system in Bangkok subsoil condition, this research
performed the consolidation analyses of piled raft foundation systems for low-rise (8-storey) and high-rise (25storey) buildings with basement levels in clay soil, using three-dimensional Finite Element Method. The soils aremodelled with Hardening Soil model and Mohr-Coulomb model. Evaluations of piled raft foundation, i.e., the load
sharing ratio of piles, settlement behaviours in the foundation system are performed. The parametric study on theeffect of raft depth, and load carried by piles of piled raft was done. The consolidation had a strong influence on theload carried by piles of the piled raft foundation in Bangkok. The load shared by piles can increase by up to 12%
and 6% for low rise and high-rise buildings, respectively due to the consolidation effect. Therefore, the design of the
piled raft foundation system in Bangkok subsoil essentially consider the consolidation effect.
Keywords: 3D-FEM, Bangkok subsoil, Consolidation, piled rafts, Piled-raft load sharing, Softs soil

referred to as piled raft foundation.
Recent years, the “Piled Raft Foundations” (PRF)
have been widely accepted as one of the mosteconomical methods of foundation systems [3, 4].
Thus, the piled raft systems have been used
extensively in many parts of the world e.g. England[5], Japan [6], Germany [7, 9, 10]. The application of
piled rafts on soft ground is becoming a significant
issue in foundation design. A few successful
applications and analysis of piled rafts on soft groundhave been reported [8, 13]. However, the behaviour of
piled raft foundation supporting the structure in clay
soil was found that the long-term effect(consolidation) increases the load carried by piles anddecreases raft contact pressure [5]. This means that theeffect of consolidation in clay has influence on piled
raft foundation system. Previous studies on numericalanalysis of piled raft foundation in Bangkok subsoil
condition indicated the potential of using PRF in
Bangkok subsoil [2, 14]. However, they considered
only short term behaviour.
To pay special to the consolidation effect, a
coupled three-dimensional (3D) mechanical andhydraulic numerical model is used to analysis the
behaviour of piled raft in Bangkok subsoil. The model
considers the dissipation of excess pore water pressure
in saturated clays. Two different building sizes, i.e.,
low-rise (8-storey) and high-rise (25-storey) buildings
with basements are considered in this study toevaluate the potential of using the piled raft system.
The main factor to be investigated its influence is thelevel of raft under long-term condition.

INTRODUCTION
In Bangkok, there are many building projects
constructed on soft soil. As the subsoil of this area
consists 13-16 m thick soft clay and stiff clay
interspersed with sand [1], the pile foundation must be
used to transfer the load to stiff soil layers. Normally,
the design and construction of foundation system onsoft ground have posed various problems togeotechnical engineers, such as consolidation,
excessive settlement, negative skin friction andbearing capacity failure. To avoid these problems, thestructure of foundation in this area are relatively
expensive.
In Thailand, the designers prefer to consider thepile group to support a structure [2]. The pile groups
mostly focus on pile capacity and group settlement
without considering the presence of the raft or mat. In
fact the foundations are built using concrete and theirbottom surfaces are attached to the soil beneath.
Therefore, in most cases end up with overdesign of thefoundation.
Typically, new office or residential buildings require 2
or more basements (depth10-20m.) for utilizing as acar park space. Meanwhile, the foundation hasconstructed in deeper level. This means that the matfoundations is placed on the stiff soil. Therefore, thesoil bearing capacity are increasing at the bottom of
the mat foundations. In recent years, the foundationengineers tend to combine these two separate systems(between shallow foundations (rafts) and deep
foundations (piles)). Such a foundation system is71

GEOMATE- Bangkok, Nov. 14-16, 2016

before consolidation.

PILED RAFT WITH CONSOLIDATION(a)

The PRF is a complex design of foundation thatcombines the bearing effect of both foundation
elements (piles and raft) [11]. In Bangkok subsoil
condition the characteristics of soft soils are highcompressibility, low shear strength and high watercontent.
Consolidation is a process by which soils decreasein volume. In general, it is the process in which
reduction in volume takes place by expulsion of water
under long term static loads. When the consolidation
settlement occurs, the soil at the bottom of the raft are
deformed. Therefore, the consolidation may have astrong influence on the load carried by raft, whichconsequently affect the load carried by piles. In thisstudy, only the settlement due to consolidation is ofinterest. The value of incremental of consolidation
settlement “∆S end ” is defined as shown in Fig 1.

(b)

Fig. 2 Reference Case (a) LR-PR, (b) HR-PR

NUMERICAL MODELING OF PILED RAFT
FOUNDATION ANALYSIS
Reference Case
A parametric study was considering 9 × 9 m, and 1
m. thick raft with 9 piles. The low-rise (8-storey) andhigh-rise (25-storey) buildings with basements are
considered in this study. Both Low-Rise-Piled Raft
(LR-PR) and High Rise-Piled Raft (HR-PR) having
identical characteristics as shown in Fig 2 (a) and 2 (b),respectively.
The pile foundation in this study was designed
following the pile group concept with a safety factor(FS) of 2.3 for each single pile, which is the current
design in engineering practice. The raft level is varied
from 0 to 10 m below the ground level. The boredpiles have 1 m diameter (d) being arranged in thefoundation with the spacing of 3 m. The level of pile
tip is at 23 (1st stiff clay layer) and 36 m (2nd sandlayer) below the ground surface for low-rise and highrise buildings, respectively. Summary of the analysis
cases is shown in Table 1.

Fig. 1 Concept of piled raft foundation withconsolidation.
Both, piles and raft are considered in the load
distribution process:(1)
Ptot= Pp + Pr
where P tot = total load of the building; P p = load
carried by the pile group; P r = load carried by the raft.
The bearing behaviour of the piled raft iscommonly described by the piled raft coefficient or
the load sharing ratio α
ofprpiles
which is defined by
the ratio between the sum of load carried by pile andthe total load of the building:α pr =∑R

α PG = 1

pile ,iRtot

Table 1.

(2)

Building(3)

where α pr
= the load sharing ratio ∑of Rpiles;
pile ,i= the amount of the pile loads; Rtot = total load of thestructure. α PG = the load sharing ratio of pile group.
(4)=
α α −α∆final

Summary of piled raft foundation ofnumerical analyses conducted.
Pile
Pile tip
Raft
Total load
spacing level
level
(kPa)
(m)
f**

Low-rise8-storey3d*

23High-rise
25-storey

3d*

36e**

initialThe effect of consolidation settlement is
considered in load sharing ratio of pile. As mentioned
α finalpreviously= the load sharing ratio of pile at end of
settlement; α intial= the load sharing ratio of pile

* d (pile diameter): 1 m.

72

(m)
0
4
8
10
0
4
8
10

140
146
152
158
350
356
362
368

GEOMATE- Bangkok, Nov. 14-16, 2016

0
-5
-10

0
D
F
JH

200

400

0

GW.

Last Piezometer
Reading (2012)

Soft clay
MediumSu
SPT

1stStiff clay
-20

Depth (m)ec LinostatiHydr

1st Sand2ndStiff clay

-35
-40

Pile tip level
low rise

ine
wn L
awdo
ic Dr
metr
Piezo

-30

-4
-8
-10

-15

-25

Su (kPa)

600 0 10 20 30 40 50

level of raft consideredin this study

Piezometric pressure (kPa)

2nd Sand-45

Pile tip levelhigh rise-50
Hard clay
-55
-60

0

200

400

600 0

20 40 60 80
SPT, NFig 3 Soil profile and Piezometric line (a), Geometry of the problem and 3D Finite element mesh used in this study (b).
drawdown pressure was considered in this study.
(b)** f: floating pile in clay; e: end bearing in sand layer.

(a)Applied Load
Uniformly Distributed Loads (UDL) is used in thisanalysis. The weight of the structure and designed
load were computed. These UDL are applied on top
surface of the raft in analyses of PRF. The basement is
considered to apply the load of 50 ton per level. The
total applied loads on each foundation are listed inTable 1.

Modeling and Boundary Condition
The geometry of the problem and FE meshsimulation of the piled raft foundation are shown in
Fig. 3(b) and Fig. 4. The 3D-FEA using PLAXIS 3D
version 2013 was carried out in this study. A coupledmechanical and hydraulic model was used for the
consolidation analysis. The 3D model included arigorous treatment of the soil and raft which wererepresented by volume elements. The piles aremodeled as embedded piles in which the pile isassumed to be a slender beam element. The boundary
conditions adopted for analyses are displacement
restraints with roller supports applied on all verticalsides and pin supports applied to the base of the mesh.
The layer surface (upper and bottom side) is allowed
to drain while the other sides are kept undrained by
imposing closed consolidation boundary conditions.

Subsoil Condition
The subsoil profile in this study are referred fromthat in the north of Bangkok. The generalized profilesof the stratified soil at the considered location are
shown in Fig 3(a). The uppermost 2.0 m thick layer isthe weathered crust, which is underlain by 6.0 m thicksoft to medium clay layer. A medium clay layer isfound at the depth of 8.0 m from the surface. Below
the medium clay is stiff clay; the thickness is about
15m. The first sand layer is generally found at a depthof 25 to 30m. Below the upper first sand layer, there isstiff clay and further down alternating layers of dense
sand and hard clay. The ground water table is below
the ground surface at 1.5 m [14, 15]. The pore waterpressure condition in Bangkok soft clay arehydrostatic from 1m below ground surface. Then thepiezometric changed to drawdown near middle of clay
layer as shown in Fig 3(a) [22]. The piezometric73

GEOMATE- Bangkok, Nov. 14-16, 2016of pile reduces from 95% to 80% and 98% to 91%
with increasing raft levels (0-10 m.) for the low-riseand high-rise buildings, respectively.
At the end of consolidation, the load sharing ratioof pile increases 2% to 12% and 1% to 6% withincreasing raft levels (0-10 m.) for the low-rise andhigh-rise buildings respectively. Significant changesof load sharing by raft are obviously observed. This
leads to the long-term load sharing by piles of 92% to98% and 97% to 99% for the low-rise and high-risebuildings respectively. This means that theconsolidation has a strong influence on the load
carried by piles of the PRF in Bangkok subsoil.

Fig 4 Adopted piled-raft geometry.

Table 3 Constitutive models and model parameters used in analyses`

Constitutive Models and Parameters
The soft clay, medium clay and first stiff clay
layers were modelled with Hardening Soil Model withsmall strain [14, 16]. The 1st -2nd sand, 2nd stiff clay
and hard clay layers were modelled with Mohr–
Coulomb model. The soil properties used in theanalyses are mainly determined from correlating local
investigated data with comprehensive in situ tests of
MRT projects [17] and previous laboratory tests from
Asian Institute of Technology (AIT) [18-20]. Table 3
summarizes the material parameters used in theanalyses.

1.00
0.98
0.96
0.94
0.92

αpr

0.90
0.88

sinitial

Raft

0.86

∆sEnd
Water

BeforeConsolidation

Water

End ofConsolidation

Pile

0.84
sinitial

0.82

Time

HR 0 mHR -4 mHR -8 m
HR-10m∆sEnd

Before Consolidation

COMPUTED RESULTS

LR 0 m
LR -4 mLR -8 mLR-10mEnd of Consolidation

0.80
0

10

20

30

40

50

Time (month)

Effect of the Load Sharing Ratio of Piles against
Time with Differential Raft Level.

Fig. 5 Load sharing ratio of piles (α pr ) versus Timeand raft level of different building types.

Fig. 5 shows the load sharing ratio of piles fordifferent raft levels below the ground surface and time
for both building types. The analysis results show thatwhen the raft was placed on deeper soil layer, the load
sharing ratio of pile has been decreased significantly.
For subsoil condition and problem characteristics inthis study, before consolidation, the load sharing ratio
74

60

GEOMATE- Bangkok, Nov. 14-16, 2016α∆/α −αInitial

PG

0.55

0.60

0.65

0.70

0.75

0

0.80

0.85
I baseGL.

0.90

II base

0.95

1.00

top stiff

Wea. Crust

Level of raft, m

2

4

LR
HR

Soft clay

6

8
Medium clay

problem characteristics in this study, the incremental
consolidation settlement “∆S end ” increases withincreasing raft level. The incremental load sharing
ratio of pile in consolidation process (α ∆ ) increase
with increasing incremental consolidation settlement∆S end . For low-rise building, the ∆S end increase from
8.6 to 10.4 mm. with increasing α ∆ from 3% to 12%
when increasing raft (0-10 m.). For the case of highrise building, the ∆S end increase from 8.6 to 10.4 mm.
with increasing α ∆ from 1% to 6.5% with increasing
raft level.
CONCLUSION

10Stiff clay

This article presents the results of numerical analysesof the PRF in the subsoil condition of north Bangkok,
using 3-D FEM to investigate the effect of raft levelon load shared by piles in Bangkok subsoil conditionand paying special attention to the consolidation effect.

Fig. 6 Variation of load sharing ratio of piles versus
raft level of different building types
Fig. 6 illustrated the computed variation of load
sharing ratio of piles versus raft level with different
building types. The variation is the difference load
sharing ratio of pile between short term and end of
consolidation process (α ∆ = α final – α initial in Eq. (4)).
The variation of load sharing ratio of piles seems todecrease with increasing raft level between -4 m to -10m.

The analysis result in terms of load shared by piles
with consolidation effect for the PRF case in this study
in Bangkok subsoil condition can be summarized asfollows:•

The consolidation had a strong influence on
the load carried by piles of piled raftfoundation in Bangkok. The load shared bypiles can increase by up to 12% and 6% for
low rise and high-rise buildings, respectively.
Therefore, the design of the piled raftfoundation system in Bangkok subsoil should
consider the consolidation.

•

The incremental consolidation settlement“∆S end ” has significant influence on theincremental load sharing ratio of piles “α ∆ ”.
The incremental load sharing ratio of pile inconsolidation process (α ∆ ) increase withincreasing
incremental
consolidation
settlement ∆S end .

α∆
0.00
0

0.02

0.04

HR 0 mHR -4 mHR -8 m
HR-10m

2

0.06

GW.

0.08

0.12

sinitial

Raft

Pile

0.14

∆sEnd
Water

BeforeConsolidation

Water

End ofConsolidation∆sEnd

sinitial

4∆SEnd, m

0.10

Time
6

LR 0 m
LR -4 mLR -8 mLR-10m

8

10
0.00

0.02

0.04

0.06

0.08

0.10

0.12

0.14

Since the pile foundation was designed using the pilegroup concept with high FS. The raft is not consideredin the design of which the FS of the pile can besmaller. Higher efficiency of the system can be
expected. Further study with less FS of pile should bedone, to confirm effectiveness of PRF.

Fig. 7 Incremental settlement at end of consolidation∆S end versus the load sharing ratio of, α ∆Settlement of PRF and Load Sharing Ratio
The incremental settlement of consolidation ∆S end
and load sharing ratio of pile between short term andend of consolidation process (α ∆ ) with different raftlevels are shown in Fig 7. The analysis results showthat the incremental consolidation settlement “∆S end ”
has significant influence on the incremental load
sharing ratio of piles “α ∆ ”. For subsoil condition and

ACKNOWLEDGEMENT
The authors gratefully acknowledge financial support
by Thailand Research Fund (TRF) and Geotechnical& Foundation Engineering Co., Ltd. (GFE) throughthe TRF-Rri Project under Contract No 5811005.
75

GEOMATE- Bangkok, Nov. 14-16, 2016

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[2] Amornfa K, Phienwej N, and Kitpayuck P,“Current practice on foundation design of highrise buildings in Bangkok, Thailand”, Lowland
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[3] Randolph M.F, “Design methods for pile groups
and piled rafts” In Proceedings of the 13th
International Conference on Soil Mechanics andFoundation Engineering, New Delhi, 1994, Vol. 5,
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[4] Poulos H. G., “Piled raft foundations: design andapplication” Geotechnique, 2001, Vol. 51(2), pp.
95-113.
[5] Hooper J.A., “Observations on the Behaviour of a
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[6] Yamashita K., Yamada T, and Kakurai M,“Simplified method for analyzing piled raftfoundations” In Proceedings of the Third
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457–464.
[7] Poulos H. G, and Davis E. H., “Pile foundation
analysis and design” New York: Wiley. 1980.
[8] Yamashita K, Kakurai M, and Yamada T,“Investigation of a piled raft foundation on stiffclay” Proceeding of the 13th International
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Engineering, New Delhi, 1994, pp. 543-546.
[9] El-Mossalamy Y, El-Nahhas F, and Essawy A,“Innovative Use of Piled Raft Foundation to
Optimize the Design of High-Rise Buildings”
10th Arab Structural Engineering Conference,
2006, pp. 13-15 November 2006, Kuwait[10] Randolph M. F., “Design of pile raft foundations”
Recent Development in Laboratory and Field
Tests and Analysis of Geotechnical Problems,
Bangkok, 1983, pp. 525-537.
[11] Katzenbach R, Arslan U and Moormann C, “Piled
raft foundation in Germany In H. J.A. (Ed.)”,Design applications of raft foundations, London,
Thomas Telford Ltd, 2000, pp. 323-391.
[12] Poulos H.G., “Piled raft and compensated piled
raft foundations for soft soil sites. Advances ondesigning and testing deep foundations” Geotech.
Spec. Publ, ASCE, 2005, Vol.129, pp. 214–235.
[13] Tan Y.C., Cheah S.W., and Taha M.R.,“Methodology for design of piled raft for 5-story
buildings on very soft clay” Foundation analysis
and design: innovative methods. Geotech, Spec.

P., “Numerical Analyses of Piled Raft Foundation
in soft soil using 3D-FEM”, South-East Asian
Geotechnical Society Journal, 2015, Vol. 46(1),pp. 109-116.
[15] Jamsawang P, Bergado D. T, and Voottipruex P,“Field behavior of stiffened deep cement mixing
piles” Ground Improvement, 2010, Vol. 164(1),pp. 33-49.
[16] Detkhong T, and Jongpradist P, “Impact ofConstitutive model on the Numerical Analysis of
Tunneling in Bangkok Clay”, Conference on
Advances in Civil Engineering for Sustainable
Development, 2014, pp. 349-354.
[17] Prust R.E., Davies J, Hu S, “Pressuremeter
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[18] Jamsawanga P, Yoobanpota N, Thanasisathita N,
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42–56[19] Bergado D.T, Youwai S, Teerawattanasuk C, and
Visudmedanukul P, “The interaction mechanism
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517-534.
[20] Tanchaisawat T, Bergado D.T., and Voottipruex,
P, “Numerical simulation and sensitivity analysesof full-scale test embankment with reinforced
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Geotextiles and Geomembranes, 2008, Vol. 26,
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[21] Sawatparnich A, Sunitsakul J, and Kanhasing P,“Numerical Analysis of DOH Bearing Unit
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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

A STUDY ON PERMEABILITY OF SILICA MICRO-PARTICLES
Kentaro Uemura1, Yukoh Hasunuma1
Takamitsu Sasaki , Koichi Nagao3, Naoaki Suemasa1 and Shunsuke Shimada4
1
Tokyo City University Graduate Division, Japan; 2Kyokado Engineering Co.,Ltd, Japan;
3
Sato Kogyo, Japan; 4 Kyokado Co.,Ltd, Japan
2

ABSTRACT
Liquefaction has drawn an attention since the Niigata earthquake occurred in 1964. Since then, many
construction methods for liquefaction countermeasures have been developed, and some of them are practicallyused at present. However, it is difficult for most of the countermeasures to be applied to the ground under existingbuildings because the large equipment necessary for doing these methods cannot be set in narrow space. In thiscase, a permeable grouting method is generally adopted for the countermeasures, in which the injection material
is required to be permeable to sandy ground and durable enough to maintain its effectiveness for long time.
However, because the cost of such injection materials is expensive, silica micro-particles were chosen as analternative injection material with relatively affordable and low cost in this study. In order to confirm thepermeation characteristics of the silica micro-particles, a series of experiments and analysis using a proposedmodel of one-dimensional injection were carried out. As a result, the characteristics of permeating and clogging
micro-particles in sand were confirmed through the experiments. In addition, the result of the analysis showed theclogging tendency similar to its experiment result
Keywords: Micro-particles, Permeability, Clogging, Liquefaction Countermeasurecan permeate even in finer sand layers. Their
durability and performance are confirmed to be as
high as those of conventional cement grout. It is,however, difficult to apply these methods to detachedhouses because these grouting materials are still
expensive. Therefore, the application of low-cost andaffordable
micro-particle
withcomparable
performance is examined in this study. One of the
improvement effects of permeating the microparticles in sand layer is to increase the density ofgrouted sandy ground. Fig.1 shows the mechanism ofground improvement by the micro-particle
permeation. According to the liquefaction evaluation
method, so called FL method, which is widely used asa simple ground liquefaction evaluation method inJapan [7], liquefaction resistance strength increases
according as fine-grained soil fraction increases. On
the contrary, several studies reported that liquefaction
resistance of sand containing non-plastic fines was
smaller than that without non-plastic fines at the same

INTRODUCTION
Not only the 2011 Tohoku earthquake but also the2016 Kumamoto earthquake caused serious
liquefaction damage to individual property. Thoughlots of researches and discussions have been done onliquefaction for long time, many existing buildings on
alluvial plains and reclaimed land are still exposed torisks of liquefaction. It is, therefore, essential todevise countermeasures against liquefaction urgently.
Many construction methods for liquefaction
countermeasures have been developed since theNiigata earthquake occurred in 1964 in Japan andsome of them are practically used at present. Most ofthese methods, however, are applied only toimportant structures and buildings because large
construction equipment is needed to perform it.
Therefore, many existing buildings with high risk ofliquefaction still remain not to be reinforced. In
general, grouting methods are applied to the groundunder such existing building because they can be
executed even in narrow space as small-facilities are
utilized for the construction and less construction
byproduct is produced. A chemical grout [1] or a veryfine cement grout [2] is widely used as an effective
injection material for the grouting method. Their longterm durability and high strength were verified
through previous laboratory tests and field
experiments, [3], [4]. In addition, recently developed
suspension grouts such as an ultrafine cement groutwith an average size of 1-3μm [5] and a superfine
spherical silica grout with sub-micrometer scale [6]Unimproved soil

Improved soil

Fig.1 Ground improvement by injected
micro-particles
77GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Physical properties of silica sandρs (g/cm3)
ρsmax (g/cm3)
ρsmin (g/cm3)

Silica sand
♯4
2.638
1.530
1.305

soil structures are so complex, the permeability of
micro-particles is still unclear [11]. In this research,in order to confirm the permeation characteristics of
micro-particles, experiments and simple analysis of
one-dimensional injection were carried out.

Silica sand
♯5
2.640
1.555
1.300

ONE-DIMENSIONAL INJECTION
EXPERIMENT

Parsentage Passing(%)

100

In this chapter, the characteristics of microparticles on permeation were confirmed by theexperiments using a short column and a multi-layer
divided column. Here, volume compressibility ofsand skeleton is neglected in these experiments. Table
1 and Fig.2 show the physical properties and the grain
size accumulation curves of silica sand ♯4 and ♯5,
respectively. A sand specimen was prepared by the
air-pluviation method at specified relative density inan acrylic column, which was saturated by injecting
de-aired water after inflow of CO2 gas. Then, a seriesof permeation tests was examined by injecting watercontaining micro-particles. The grain size
accumulation curve of the micro-particles is shown in
Fig.3.

80
60
40
20

Silica Sand ♯4
Silica Sand ♯5

0

0.1

1

10

Grain size(mm)Fig.2 Grain size accumulation curve ofsilica sand #4 and #5

Parsentage Passing(%)

100
80

Experiments Using Short Columns
60

Conditions of Experiment40

A series of injection laboratory experiments was
conducted in order to confirm the permeability of
micro-particles which way vary with the conditionsof sand specimen and the initial injection velocity
permeated into. In this test, silica sands ♯4 and ♯5
were used for the specimens and their physicalproperties and grain size accumulation curves areshown in Table 1 and Fig.2, respectively. The sandspecimen with 5.0cm in diameter and 15.0cm tall was
prepared by the air-pluviation method at relative
density of 60%. The injection material mixed withwater, which weight concentration of micro-particle
in the mixed water was set to 4.0%, was permeatedinto the specimen. The volume of the injection
material accumulated in the specimen was estimatedby measuring the weight change of the specimen witha platform scale. Table 2 shows the conditions ofthese experiments. Different grouting velocity was set
to each specimen by changing injection pressure. In
this experiment, Injection pressure controlled so as
not to be too high. And also, the concentration of thedrained injection material obtained at eachexperiment was measured by evaporating in a drying
furnace.

20
Micro-particle
0
0.001

0.01

0.1

1

Grain Size(mm)Fig.3 Grain size accumulation curveof micro-particles
Table 2 Conditions of experiment
Case
Case1
Case2
Case3
Case4
Case5
Case6

Silica
sand
♯4
♯4
♯4
♯5
♯5
♯5

Concentrationof micro-particle
4.0%
4.0%
4.0%
4.0%
4.0%
4.0%

Injectionpressure
10 kPa
30 kPa60 kPa
10 kPa
30 kPa60 kPadensity [8], [9], [10]. At this point, it is not clear inwhat way micro-particles are permeated anddeposited in sand layer and whether the depositedparticles increase the liquefaction resistance of thesand. In particular, as ground layer conditions and the78

GEOMATE- Bangkok, Nov. 14-16, 2016

Volume of injected mixture(cm3)

Results of ExperimentFig.4 to Fig.6 show the results of permeation
experiments. Broken lines in Fig.4 and Fig.5 indicate
the approximate lines obtained from the rate of
injection volume observed immediately after the startof injection. The results of Case1 to Case3 are shownin Fig.4, which indicates the change of the injection
volume of water mixed with micro-particles withelapsed time. It is shown that the injection volume
increased linearly with time in Case1, Case2 andCase3. Accordingly, it indicates that the clogging of
the injected micro-particles did not occur at the poreof silica sand ♯4.
The results of Case4 to Case6 are shown in Fig.5.
The injection volume increased linearly in Case5 andCase6. On the other hand, the rate of injection volume
gradually decreased in Case4 whose injection
velocity was set at the lowest of the three Cases. This
result indicates that the clogging of permeated microparticles was likely to be dependent on the velocity ofpermeation. In this experiment, however, it was notclear in which part of sand specimen the clogging
occurred.
Fig.6 shows the change of weight concentration of
injected micro-particles containing in the draindischarged from specimen. From the results that theconcentrations of micro-particles in the drainage wereconstant at the initial value in all cases except for
Case4, it was inferred that the micro-particles
permeated uniformly in the sand specimen. In
contrast, it was confirmed that the concentration in
the drainage reduced with elapsed time in Case 4.

2000 yy = 28.2x

1500

y = 19.2xy = 9.9x

1000
Case1
Case2
Case3

500
0
0

40

80

120

x

160

Time (s)Volume of injected mixture(cm3)

Fig.4 The result of Case1 to Case3
2000 yy = 19.0x

Case4
Case5

1500

y = 12.2x

Case6
1000

y = 8.3x

500
x

0
0

40

80

120

160

Time (s)Fig.5 The result of Case4 to Case6
Concentration of
micro-particles in drainage
(%)

5.0
4.5
Case1
Case2
Case3
Case4
Case5
Case6

4.0
3.5

3.0
2.5
2.0
0

50

100

150

200

Time (s)Fig.6 Concentration of micro-particles indrainage
Table 3 Conditions of experiment

Experiment Using a Multi-layer Divided Column
Conditions of Experiment

Case

In order to obtain the clogging mechanism of
micro-particles, it is necessary to disclose in whichpart of sand specimen clogging occurred.
Experiments using a multi-layer divided column were
performed to clarify the distribution of injected
micro-particles concentration.
The multi-layer divided column made fromacrylic resin are 5.0cm in diameter and 50.0cm talland separable by 10.0cm. The experimental apparatus
is shown in Fig.7. In this test, silica sand ♯5 was used
for sand specimens, which physical properties andgrain size accumulation curve are shown in Table 1
and Fig.2, respectively. The sand specimen was
prepared in the column by the air-pluviation method

Case7
Case8
Case9
Case10
Case11

Silica
sand
♯5
♯5
♯5
♯5
♯5

Concentrationof mixture
4.0%
4.0%
4.0%
6.0%
6.0%

Injection
presure10 kPa
20 kPa
40 kPa
10 kPa
40 kPaat a relative density of 60%. The weight concentrationof micro-particles in injected water was set to 4.0%
or 6.0%, and the mixture was permeated into thespecimen. In order to reveal difference of
permeability due to initial injection velocity, different
grouting velocity was set to each specimen bychanging injection pressure. The conditions ofexperiments are shown in Table 3. The column was
79GEOMATE- Bangkok, Nov. 14-16, 2016

Volume of injected mixture(cm3)

divided after the injection and then the concentrationof micro-particles in pore water was measured byevaporating in a drying furnace. The concentration of
the micro-particles measured was plotted at the centerof the depth of each divided piece on graphs.
Mixture of micro-particle and waterAirpressure

y=10.2x1500

y=5.8x

1000

500
x

0
0

100

200
300
Time(s)400500(a) Results of Case7 to case9

Specimen
Volume of injected mixture(cm3)

50.0cmPlatform
scale

Case7
Case8
Case9

y=13.6x2000

10.0cmDrainage2500 yFig.7 Experimental apparatus
Results of experiment2500 yCase10
Case112000

y = 13.3x

1500

y = 4.9x

1000
500

x

0
0

The results of experiments are shown in Fig.8 andFig.9. Fig.8 are the relationship between the changeof the injection volume of mixture with time. Fig.9
are the relationship between micro-particles indrainage and permeation distance. Fig.8 shows thatthe concentration of injected micro-particles indrainage was constant in Case9 whose velocity was
the highest of the three Cases. It indicates that the
injected micro-particles flowed through the pore of
the specimen without clogging. In Case7 and Case8,
on the other hand, the concentration in drainage
tended to be higher in the order of shorter distancefrom the injection source. It indicates that clogging
tended to occur at a short distance from the injection
point in Case7 and Case8 whose velocities werelower than Case9.
The results of Relationship between microparticles in drainage and permeation distance areshown in Fig.9. The comparison between Case7 andCase10 revealed that the highest concentration in
Case10 was larger than that in Case7. In addition, theconcentration in Case10 tended to develop
significantly near the injection source. Therefore, it
was also presumed that the trend of clogging depends
on the concentration of injected micro-particles.
From these results, it was confirmed that clogging
was affected by the properties of sand and microparticles, injection velocity, injection volume and theconcentration of micro-particles. Particularly, in thecases whose injection velocity was low, since themicro-particles clogged in the sand layer near the
injection source, the rate of injection volume became
low.

100

200
300
Time(s)400500(b) Results of Case10 to Case11
Fig.8 The change of injection volume of
mixture with time

Concentration of
micro-particles in drainage
(%)

10.0
Case7
Case8
Case9

8.0
6.0

4.0
2.0
0.0

0

10
20
30
40
Permeation distance (cm)50

Concentration of
micro-particles in drainage(%)(a) Results of Case7 to case9
10.0

Case10
Case11

8.0
6.0
4.0
2.0
0.0
0

10
20
30
40
Permeation distance (cm)50(b) Results of Case10 to Case11
Fig.9 Relationship between micro-particles indrainage and permeation distance80

GEOMATE- Bangkok, Nov. 14-16, 2016

ANALYSIS OF ONE-DIMENSIONALPERMEATION(divide into n elements)
C0

C0

Conditions of Analysis

Cs1q

One-dimensional permeation analysis was
performed to understand the influence of clogging by
micro-particles. A model was created based on the
results from the previous one-dimensional injection
experiments. Here, volume compressibility of sandskeleton was neglected in this chapter. Simulation
was carried out under the same conditions as theinjection experiments. The analytical model is shownin Fig.10. The analysis object region was 50.0cm talland divided into 100 elements. A permeability
coefficient was determined depending on the amountof micro-particles storage in each divided element. In
the analysis, the concentration of permeated microparticles is expressed as the volumetric concentration
C, which is equal to the initial concentration of microparticles C0 at the boundary of injection source anddecreases with permeation distance depending on theamount of micro-particles storage.
From the experimental results, it was revealed thatthe amount of micro-particles storage increased whensome of the permeated micro-particles were caught inthe void of sand skeleton. Therefore, theconcentration of micro-particles, some of which are
adhered in the pore of sand and others of which areadvected, is formulated by the following advection
diffusion equation (1).
𝜕𝐶
𝜕𝑡

+υ

𝜕𝐶
𝜕𝑥

=−

𝜕𝐶𝑠
𝜕𝑡

A

Cs2

k2

C2
Ci

50cm
Csi

ki

Initial element

Ci+1

Filled element
Cn-1
Csn-1
Cn

kn-1
Cn

Fig.10 Analytical model
Table 4 Condition of analysis
Case

Case7
Case8
Case9

Parameter
a
b
4.00
0.07
2.20
0.04
0.40
0.01

the pass length. According to Eq. (2), the more theamount of micro-particle in an element became, thesmaller permeability coefficient tended to be. As thevolume of injected water mixture increases, theamount of accumulated micro-particles increases insand void. The accumulation of the micro-particles isexpressed in Eq. (3).

(1)

where C : the volumetric concentration of microparticles in injected mixture, Cs : the adhesion
concentration obtained by dividing the volume ofaccumulated micro-particles in an element with thepore volume of the element; x : the permeation
distance from the injection source; t : time; 𝜐 : theseepage velocity. From the results of the experiments,
it was suggested that clogging of micro-particles wasaffected by injection volume, injection velocity andthe concentration of micro-particles. Therefore, in theanalysis, the clogging was simulated by using the passlength I which means how many micro-particles
passed through a boundary surface. The pass length
was obtained by dividing the total volume of injectedwater mixture flowing through a boundary surface
with the area of the surface. Moreover, assuming thatthe seepage velocity of water mixture follows theDarcy’s low, a permeability coefficient ke for eachelement which is reduced by clogging is defined as
Eq. (2).
𝑘𝑒 = 𝑘0 𝑒 −𝑎･𝐼

k1
C1

𝐶𝑠 = 𝐶𝑠𝑚𝑎𝑥 × tanh⁡(𝑏･𝐼)(3)where Csmax: the limit accumulation volume of microparticle, b : the accumulation parameter. In Eq. (4),the accumulation of micro-particles in an element
increased by increasing the amount of watercontaining micro-particles that flowed through theelement. The experiments of Case7 to Case9 using themulti-layer divided column were simulated in thisanalysis. The parameters used in this simulation are
shown in Table 4.
Results of Analysis
Analysis results are shown in Fig.11. In Fig.11 (b),C and C0 are the concentration of micro-particles inan element and the initial concentration in mixture,
respectively. Moreover, the plots indicate the resultsof experiments and the broken lines that of analysis.
From all cases, it was confirmed that the results ofexperiments and analysis closely matched each other.
Particularly in Case7 and Case8, non-linear behavior(2)

where ke : the reduced coefficient of permeability byclogging, k0 : the initial coefficient of permeability,
e :Nepier’s constant, a : the parameter of clogging, I :81

GEOMATE- Bangkok, Nov. 14-16, 2016

was observed in the concentration rate C/C0 and larger
value was seen near the injection source.
It is considered that the results of experiments
were able to be simulated by the proposed model.
However, an interrelationship among the parameters
is unclear. To solve this problem, further studies are
required.

REFERENCE[1] Yonekura R, Shimada S, “Long Term Durabilityof Chemical Grouts”, Tsuchi-to-Kiso, Vol.40
No.12, Japanese Geotechnical Society, 1992, pp.
17–22. (In Japanese)[2] Zebovltz S, Krizek R J, Atmatzidis D K,
“Injection of Fine Sands with Very Fine Cement
Grout”, Journal of Geotechnical Engineering,
Vol.115 No.12, 1989, pp. 1717-1733[3] Nakayama T, Sawada R, Hiraoka Y, Akagi H,“Evaluation of Long Term Stability ofChemically Stabilized Sand with Grouting”,
J.JSCE Ser.C Geotechnical Engineering, Vol.69
No.2, 2013, pp. 162-173. (In Japanese)[4] Fukushima K, Kwachi T, “Evaluation about
Durability of Soil Grouted with Metal-Acrylate
on Site”, Proceedings of the 33th Annual
Conference of the Japanese Geotechnical Society,1998, pp. 2181-2182. (In Japanese)[5] Koizumi Y, Tanaka T, Takeuchi J, Kanazawa T,
Nishigaki M, “Development of Grouting Methodof Fine Sand with Ultrafine Cement Grout”,
Journal of the Society of Materials Science,
Vol.61 No.1, 2012, pp. 52-57. (In Japanese)[6] Yamada T, Sekine I, Sekiguchi T, Ishida H, Seki
H, Fukushima N, “The Project for Grouting
Development for High Level Radioactive Waste
Repository, 4; Development of Superfine
Spherical Silica Grout”, Proceedings of the 65th
Annual Conference of the Japan Society of Civil
Engineers, 2010, pp. 89-90, (In Japanese)[7] Architectural
InstituteofJapan,
Recommendations for Design of Ground
Improvement for Building Foundations:
Maruzen.Inc, 2008, pp. 313-342. (In Japanese)[8] Polito C P, Martin J R, ”Effects of Nonplastic
Fines on the Liquefaction Resistance of Sands”,
Journal of Geotechnical and Geoenvironmental
Engineering, 2001, pp. 408-415.
[9] Kuwano J, Sapkota B K,Hasizume H, Takahara
K, “Liquefaction Characteristics of Sand
Containing Fines”, Tsuchi-to-Kiso, Vol.47 No.7,
The Japanese Geotechnical Society, 1993, pp.
23-28–22. (In Japanese)[10] Sato M, Oda M, Kazama H Ozeki K,“Fundamental Study on the Effect of Fines on
Liquefaction Strength of Reclaimed Ground”,
J.JSCE Ser.III Geotechnical Engineering,
No.561, 1997, pp. 271-282. (In Japanese)[11] Yoneda S, Nakagawa K, “Basic Properties andPenetration of Cement Suspension Grouts withDifferent Grain Sizes”, J.JSCE Ser.IV
Infrastructure Planning and Management,
No.462, 1993, pp. 101-110. (In Japanese)

Volume of injected mixture (cm3)

2000
Case7(experimental)
Case8(experimental)
Case9(experimental)
Case7(analysis)
Case8(analysis)
Case9(analysis)

1600

1200

800

400

0
0

50

100

150

200

250

300

Time (s)(a) Relationship volume of
injected mixture and time
2.0

C/C0

1.5

1.0

0.5

Case7(experimantal)

Case7(analysis)

Case8(experimental)

Case8(analysis)

Case9(experimental)

Case9(anakysis)

0.0
0

10

20

30

40

50

Permeation distance(cm)(b) Relationship C/C0 and permeation distanceFig.11 Results of analysis
CONCLUSION
Experiments using a short column and a multilayer divided column were carried out to confirm thecharacteristics of micro-particles permeating andclogging. Permeability of micro-particles was
influenced by seepage velocity, injection volume andthe concentration of micro-particles. In particular, if
the seepage velocity of injected water mixture washigh, clogging scarcely occurred because of its
sweeping action. On the other hand, when the seepage
velocity of the mixture was low, decreasing thesweeping action occurred the clogging of microparticles. From the results of experiments, it was also
revealed that clogging of micro-particles significantly
grew near the injection point.
An one-dimensional permeation analysis was
performed to understand the influence of clogging of
micro-particles. It was confirmed that the results ofexperiments and analysis closely matched each other.
However, further studies are required because aninterrelationship among the parameters is unclear.

82

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DOUBLY ASYMPTOTIC OPEN BOUNDARY CONDITION FOR
MODAL RESPONSES OF PORE WATER PRESSURE
Suriyon Prempramote
Department of Civil Engineering, Kasetsart University, Thailand

ABSTRACTThe doubly asymptotic open boundary condition has been developed for solving the modal transient seepage
equation of any isotropically saturated soil layer with a constant depth. The continued fraction technique isadopted to formulate the doubly asymptotic continued fraction solution in the frequency domain. All constants ofthe solution are determined at both high- and low-frequency limits. The convergence rate of the solution is much
faster in comparison with the high-frequency continued fraction solution. By introducing the auxiliary variables
and the doubly asymptotic continued fraction solution to the relationship between the modal seepage flow andthe modal response of pore water pressure, the open boundary condition is constructed first in the frequency
domain. After transformed to the time domain, the open boundary condition is in terms of a system of fractional
differential equations that can be solved by the direct integration methods. The convolution integral is no longer
required. The results of modal responses of pore water pressure obtained from the doubly asymptotic open
boundary condition with merely low orders of continued fraction are very accurate.
Keywords: Seepage, Open boundary condition, Continued fraction, Modal dynamic permeability, Doubly
asymptotic

an interesting one proposed in terms of a highfrequency open boundary condition of transient
seepage equation for semi-infinite layers with aconstant depth [10]. The high-frequency open
boundary condition is expressed as a system offractional differential equations in the time domainthat can be solved by the Newmark method with the
improved numerical method of Riemann-Liouvillefractional derivative [11]. The concept of
formulation of this open boundary condition is fromthe open boundary condition of scalar wave equationin [6].
Recently, a high-frequency open boundary for
two-dimensional transient seepage problems insaturated soil layers with a constant depth has beensuccessfully developed by extending the scaled
boundary finite element method (SBFEM) [12] withthe concept used in [10]. Under either a short-termflow or a long-term flow, the responses of porewater pressure obtained from the high-frequency
open boundary are very accurate as long as the orderof continued fraction chosen by the user is highenough. Even though it provides accurate results of
responses, it may be inefficient if a very high orderof continued fraction is chosen.
Thus, the objective of this paper is aimed atpresenting the doubly asymptotic open boundary
condition, which is more efficient, for modal
responses of pore water pressure in saturated soil
layers with a constant depth. The convergence rateof the solution obtained from this open boundary
condition is much faster, and only a few orders of

INTRODUCTION
One of the important issues in simulation ofwave propagation in an elastodynamic system is themodeling of the unbounded domain e.g. foundationsoil [1]. The responses of the structure on theunbounded domain are often affected by the size ofthe unbounded domain used in the analysis. If thesize is very large, the analysis results of responses
are more accurate, but if the size is not large enough,the results may be polluted by the spurious reflection
at the truncated boundary of the unbounded domain[2]. This is a common issue when using the finite
element method (FEM) as a tool in modeling.
To cope with this issue, a non-reflecting
boundary condition have to be imposed on thetruncated boundary to satisfy the boundary condition
at infinity, in other words, the radiation condition [3].
This non-reflecting boundary condition is actuallyan approximation based on mathematicalformulations. Today, there are several non-reflecting
boundary conditions called by different names e.g.
artificial boundary condition [4], absorbing
boundary condition [5], open boundary condition [6],etc.
Besides the responses of the structure, theresponses of the unbounded domain are also
important, especially the response of pore waterpressure that may significantly affect the embedded
part of the structure in transient seepage problems.
Many approaches have been proposed to solve suchproblems [7-9]. Apart from such approaches, there is83

GEOMATE- Bangkok, Nov. 14-16, 2016

where the introduced a 0 denotes the dimensionlessfrequency, which is defined ascontinued fraction are required to obtain an accurate
result.
MODAL DYNAMIC PERMEABILITY
COEFFICIENT
For any isotropically saturated soil layer with aconstant depth h underlain by an impervious layer asshown in Fig. (1), the governing equation of twodimensional transient seepage in the time domain isgiven by

a0 =

ωh 2
c

(5)

Both sides of Eq. (4) must be equal to the same
constant which is denoted as λ2. This allows Eq. (4)to be separated into the following equations:~λ ~Z , zz +  Z = 0
h
(1)
2

1
∇ 2u = uc1~~
X , xx − 2 ((ia0 ) + λ2 ) X = 0hwhere u = u(x,z,t) is the response of pore waterpressure and c is the hydraulic diffusivity, which isequal to the ratio of the permeability k to the specificstorage S s . Note that the arguments of the function u
in Eq. (1) are omitted for simplicity in thenomenclature. The initial and boundary conditions
are given by u̇ (0,z,0) = 0 at the vertical boundary Γ V .
At the upper boundary Γ U , u(x,0,t) = 0. And at thelower boundary Γ L , which is adjacent to theimpervious layer, u(x,h,t), z = 0.

(6)

(7)

The solution of Eq. (6) is in the form of

~
λ 
λ 
Z = C1 cos z  + C 2 sin  z 
h 
h 

(8)

where C 1 and C 2 are constants that can be
determined from substituting the initial condition Z�= 0 at z = 0 into Eq. (8). The constant C 1 is found tobe zero while C 2 is arbitrary and negligible because
it never be calculated. As a result, Eq. (8) is reduced
to be the eigenfunction,

 λj 
~Z j = sin  z 
h 

(9)

where the eigenvalue is determined from
Fig. 1

Saturated soil layer with constant depth.

Equation (1) can be rewritten in the frequency
domain as

1
U , xx +U , zz − (iω )U = 0cwhere U = U(x,z,ω) is the amplitude of pore waterpressure, i = √−1 is the imaginary unit, and ω
denotes the excitation frequency. The solution of Eq.
(2) is determined from the method of separation ofvariables as expressed in the following equation:

~~
U=XZ
�=X
�(x,ω) and Z� = Z� (z). Substituting Eq. (3)where X
into Eq. (2) and multiplying the equation by h2 yield

~
~
X,
Z,
h 2 ~xx − (ia0 ) = −h 2 ~zz
X
Z

λj =

(2 j + 1)π2(10)

for each mode number i.e. j = 1, 2, 3, …, and so on.
This implies that Eq. (7) must depend on λ j as well.
(2) Therefore, Eq. (7) can be rewritten in the frequency
domain for each mode as

1~~
X j , xx − 2 ((ia0 ) + λ2j ) X j = 0h(11)which is the modal seepage equation in the xdirection. The solution of Eq. (11) can be
determined from the relationship between the modal
� and the(3) amplitude of prescribed seepage flow Qj�j on themodal amplitude of pore water pressure X
vertical boundary Γ V as expressed in the following
equation:~~∞ ~(4) Q j = S j X j

84

(12)

GEOMATE- Bangkok, Nov. 14-16, 2016∞∞

where S� j = S� j (a0 ) denotes the modal dynamic
� (x,a0 ) is equal
� =Q
permeability coefficient, and Q
j
j
to

~
~
Q j = −hX j , x~~~
S j∞ = (ia0 )1 / 2 C∞ , j − λ2j (Y j(1) ) −1

(19)

with the recurrence relation(13) Y~ (i ) = (ia )1 / 2 Y~ (i ) − λ2 (Y~ (i +1) ) −1

(20)

j

Substituting Eq. (13) into Eq. (12) and rearrange theequation as

0

1, j

j

j

(i)
� ∞,j and Y� 1,j
arefor i = 1, 2, 3, …, M H where C
2 � (1) −1
2 � (i+1) −1
constants while λj (Y )and λj (Y
)
are
j

1~ ~
~
X j , x = − S j∞ X jhDifferentiating Eq. (14) with respect to x andsubstituting Eq. (14) into the differentiated equation
lead to1 ~~~X j , xx = 2 ( S j∞ ) 2 X jhSubstituting Eq. (15) into Eq. (11) and eliminating�j in the equation lead to the modal dynamicX
permeability equation,~( S j∞ ) 2 − λ2j − (ia0 ) = 0

j(14) residual terms. Substituting Eq. (19) into Eq. (16)leads to an equation of a power series (ia 0 ) including
the following terms:~~~(ia0 )(C∞2 , j − 1) + (−λ2j − (ia0 )1 / 2 2C∞2 , j λ2j (Y j(1) ) −1~
+ λ4j (Y j(1) ) −2 ) = 0
(15)

(21)

Setting the 1st term ((ia 0 ) term) equal to zero leads to~
C∞ , j = 1(22)

� ∞,j is chosen to satisfy
Note that the positive root of C
the boundary condition at infinity. Multiplying the(16)(1)
last term by −(Y� )2 /λ2j and substituting Eq. (22)j

The exact solution of Eq. (16) is of the positive form~
S j∞ = λ2j + (ia0 )into it result in

(17)

(23)

which is a residual equation with the introduced
constant

to satisfy the boundary condition at infinity. On thevertical boundary Γ V , the response of modal porewater pressure x� j = x� j (t) due to prescribed modal
seepage flow q� j = q� j (t) in the time domain can be
determined from the following convolution integral:~x = τ (~s ∞ ) −1q~ dτ
j
∫ j j

~
~
(Y j(i ) ) 2 + (ia0 )1 / 2 b1(,i j)Y j(i ) − λ2j = 0

b1(,i j) = 2(24)

for the case of i = 1. If substitute Eq. (20) into Eq.
(23), the following equation:(18)

where (s̃ ∞j )−1 = (s̃ ∞j (t−τ))−1 is the inverse Fourier~~(ia0 )((Y1,(ji ) ) 2 + b1(,ij)Y1,(ji ) ) + (−λ2j − (ia0 )1 / 2 λ2j (b1(,ij) +~ ~~2Y1,(ij) )(Y j(1) ) −1 + λ4j (Y j(i +1) ) −2 ) = 0

DOUBLY ASYMPTOTIC CONTINUEDFRACTION SOLUTION

which is also a power series of (ia 0 ) is obtained.
Again, set the 1st term ((ia 0 ) term) equal to zero to
obtain0∞ −1transform of (S� j ) .

~Y1,(ji ) = −b1(,ij)

In the derivation of the doubly asymptotic
continued fraction solution, the high-frequency
continued fraction solution is derived first and thenfollowed by the low-frequency continued fraction
solution.

(25)

(26)

)2 /λ2j andThen, multiply the last term by −(Y� jsubstituting Eq. (26) into it to obtain the recursive
equation,(i+1)

High-frequency Continued Fraction Solution~~(Y j(i +1) ) 2 + (ia0 )1 / 2 b1(,i j+1)Y j(i +1) − λ2j = 0

The continued fraction solution at the highfrequency limit (a 0 → ∞) is defined as
85

(27)

GEOMATE- Bangkok, Nov. 14-16, 2016~~
(YL(,ij) ) 2 + bL(i0), jYL(,ij) − (ia0 ) = 0

with the introduced constantb1(,ij+1) = −b1(,ij)(28)

For simplicity in notation, Y� j H and b1,j H
are rewritten as Y� L,j and bL1,j , respectively. Thus, Eq.
(27) with i = M H can be rewritten as(37)(M +1)~~(YL , j ) 2 + (ia0 )1/ 2 bL1, jYL , j − λ2j = 0

for the case of i = 1. If substitute Eq. (31) into Eq.
(36), the following equation:~~~((YL(0i ,) j ) 2 + bL(i0), jYL(0i ,) j ) − (ia0 )(1 + bL(i0), j (YL(,ij+1) ) −1 +(29) ~~~
2YL(0i ), j (YL(,ij+1) ) −1 − (ia0 )(YL(,ij+1) ) −2 ) = 0

The continued fraction solution at the low-frequency
limit (a 0 → 0) is defined as~~~~
YL , j = YL 0, j + (ia0 )1 / 2 YL1, j − (ia0 )(YL(,1j) ) −1

(30)

(38)

which is also a power series of (ia 0 ) is obtained.
Again, set the 1st term (constant term) equal to zero
to obtain~
YL(0i ), j = −bL(i0), j

with the recurrence relation(39)(31) Then, multiply the last term by (Y� L,j )2 andsubstituting Eq. (39) into it to obtain the recursive
equation,
(i+1)

~
~
~
YL(,ij) = YL(0i ), j − (ia0 )(YL(,ij+1) ) −1

for i = 1, 2, 3, …, M L where Y� L0,j , Y� L1,j and Y� L0,j are(1)(i+1) −1
) areconstants while (ia0 )(Y� )−1 and (ia0 )(Y�
(i)

L,j

which is a residual equation with the introduced
constant~bL(i0), j = 2YL 0, j

Low-frequency Continued Fraction Solution(M +1)(36)~~(YL(,ij+1) ) 2 + bL(i0+,1j)YL(,ij+1) − (ia0 ) = 0

L,jresidual terms. Substituting Eq. (30) into Eq. (29)leads to an equation of a power series (ia 0 ) including
the following terms:~~ ~~(YL20, j − λ2j ) + (ia0 )1 / 2 (2YL 0, jYL1, j + bL1, jYL 0, j ) + (ia0 )(~~~ ~(YL21, j + bL1, jYL1, j ) − 2YL 0, j (YL(,1j) ) −1 − (ia0 )1 / 2 (bL1, j +~~~2YL1, j )(YL(,1j) ) −1 + (ia0 )(YL(,1j) ) −2 ) = 0

which is a residual equation with the introduced
constantbL(i0+,1j) = −bL(i0), j(32)

This equation is satisfied by setting all terms equal
to zero. The 1st term (constant term) leads to~
YL 0, j = ±λ j
In Eq. (33), if M H is an odd number, the positive
sign is chosen. On the contrary, the negative sign ischosen when M H is an even number. The 2nd term((ia 0 )1/2 term) leads to~
YL1, j = −bL1, j / 2(41)

The doubly asymptotic continued fraction
solution, for example, with M H = M L = 1 can beconstructed by using Eqs. (19), (20), (30) and (31) as
expressed in the following equation:~∞
1/ 2 ~(33) S j = (ia0 ) C∞ , j −λ2j~(ia0 )1/ 2 Y1,(1j ) −

λ2j
(ia )
~
~
YL 0, j + (ia0 )1/ 2 YL1, j − ~ (10)
YL 0, j(42)
Note that the residual term (ia0 )(Y� L,j )−1 in Eq. (31)(34) is approximated as zero when i = M L .
(i+1)

The last term ((ia 0 ) term) can be reduced to~ ~~− 1 − 2YL 0, j (YL(,1j) ) −1 + (ia0 )(YL(,1j) ) −2 = 0(40)

DOUBLY ASYMPTOTIC OPEN BOUNDARYCONDITION(35)

after Eq. (34) is substituted into it. Then, multiply(1)
Eq. (35) by −(Y� L,j )2 to obtain86

To construct the doubly asymptotic open
boundary condition in the frequency domain, first
substitute Eq. (19) into Eq. (12) and rearrange theequation to obtain

GEOMATE- Bangkok, Nov. 14-16, 2016

~ ~~~(ia0 )1/ 2 C∞ , j X j − λ j X (j1) = Q j

~

~

~

(43) X L( 2, j) = (ia0 )1/ 2 (YL(,2j) ) −1 X L(1,)j�(1)where the introduced auxiliary variable X
j is

~
~
~
X (j1) = λ j (Y j(1) ) −1 X j(44)Then, substitute Eq. (20) with i = 1 into Eq. (44) andrearrange the equation as(53)

Again, if the same procedure done previously isrepeated for i = 2, 3,…, M L , equations with the same
pattern as Eq. (52), that is,~~ ~~− (ia0 )1 / 2 X L(i,−j1) + YL(0i ), j X L(i,)j − (ia0 )1 / 2 X L(i,+j1) = 0(54)

for i = 2, 3,…, M L are also obtained. Note that the(i+1) −1
residual term (ia0 )(Y�)in Eq. (31) is
L,j

~
~ ~
~
− λ j X j + (ia0 )1 / 2 Y1,(1j ) X (j1) − λ j X (j 2) = 0
�j is
where the introduced auxiliary variable X
(2)

~
~
~
X (j 2) = λ j (Y j( 2) ) −1 X (j1)(45) approximated as zero when i = M L . Then, combine
Eqs. (43), (45), (47), (48), (50), (52), (54) and use
Eq. (5) to form the following equation in matrix
form:
1/ 2
(46) ([ K h ] + (iω ) [Ch ]){ X } = {Q}(1)(ML ) T
�j , X
�(1)�(MH ) , X
�L,j , X
�L,j�L,jwhere {X} = [X
,…, X
] ,
j ,…, Xj� , 0,…, 0, 0, 0,…, 0] T and the time{Q} = [Qjindependent matrices,

Again, if the same procedure done previously isrepeated for i = 2, 3,…, M H , equations with the same
pattern as Eq. (45), that is,~~ ~~− λ j X (ji−1) + (ia0 )1 / 2 Y1,(ji ) X (ji ) − λ j X (ji+1) = 0(47)

for i = 2, 3,…, M H are also obtained. For simplicity
H +1)
�(M
�L,j . Thus, Eq.
in notation, X
is rewritten as Xj(47) with i = M H can be rewritten as~~~~− λ j X (j M H −1) + (ia0 )1/ 2 Y1,(jM H ) X (j M H ) − λ j X L , j = 0
�L,j iswhere the introduced auxiliary variable X~~~
X L , j = λ j (YL , j ) −1 X (j M H )Next, substitute Eq. (30) into Eq. (49) and rearrange
the equation as

~
~ ~
~ ~
− λ j X (j M H ) + YL 0, j X L , j + (ia0 )1/ 2 YL1, j X L , j~− (ia0 )1/ 2 X L(1,)j = 0
�L,j iswhere the introduced auxiliary variable X


− λ
 j


[Kh ] = 




(48)



�L,j iswhere the introduced auxiliary variable X
(2)

− λj
− λj



− λj

− λj
~
YL 0, j~
YL(01,) j












~(ML ) 
YL 0, j 

(56)

(57)

Equation (55) can be transformed into the time
domain as(51) [ K h ]{x} + [C h ]D p {x} = {q}(58)which is the doubly asymptotic open boundary
condition for solving the modal responses of pore
where
water
pressure𝑥𝑥�jNext, substitute Eq. (31) with i = 1 into Eq. (51) andrearrange the equation as~~ ~~− (ia0 )1 / 2 X L , j + YL(01,) j X L(1,)j − (ia0 )1 / 2 X L( 2, j) = 0

− λj

~

C∞ , j


~ (1)
Y1, j
(49)







~(MH )Y1, jh [Ch ] =~
c
YL1, j − 1



−1





1
−




−1
(50)

(1)

~
~
~
X L(1,)j = (ia0 )1/ 2 (YL(,1j) ) −1 X L , j

(55)

(1)

(M )

(1)

(M )

and
(52) {x} = [𝑥𝑥�j , x� j ,…, x� j H , x� L,j , x� L,j ,…, x� L,j L ]TpT{q} = [q� j , 0,…, 0, 0, 0,…, 0] , and D denotes thefractional derivative of an order p which is equal to
1/2. The vector {x} in Eq. (58) can be solved by the
direct integration methods e.g. the Newmark method87

GEOMATE- Bangkok, Nov. 14-16, 2016

with γ = 0.5 and β = 0.25 (the average acceleration
scheme) in association with the improved numericalmethod of Riemann–Liouville fractional derivative[11].
NUMERICAL EXAMPLES
Two soil layers are chosen for the analyses in thefrequency and time domains. The 1st soil layer is of
the following properties: h = 1 m and c = 1 m2/s,
while the 2nd soil layer is assumed to be deeper andless permeable, that is, h = 10 m and c = 0.01 m2/s.

The accuracy of the doubly asymptotic continued
fraction solution obtained from using Eqs. (19), (20),(30) and (31) is evaluated and compared with that of
the high-frequency continued fraction solution
proposed in [10]. The exact solution (Eq. (17)) is
used for verifying the accuracy of both solutions.
For the 1st soil layer, the doubly asymptotic
continued fraction solution with M H = M L = 1 istested at the high mode j of 1,000 with λ j =2,001(π/2). The result normalized by λ j is plotted inFig. 2(a) for the real part and Fig. 2(b) for theimaginary part. It is clear that both real part andimaginary part match those of the exact solution
throughout the entire range of a 0 . In comparison
with the high-frequency continued fraction solution
with the same number of fractions i.e. M H = 3, theyboth yield the same result. However, when tested at
the lowest mode j of 0 with λ j = π/2, the highfrequency continued fraction solution with M H = 3 isvery inaccurate (see Fig. 3). In order to gain theaccurate result, the order M H must increase to 19. In
contrast, the doubly asymptotic continued fraction
solution with M H = M L = 1 is still accurate.

(b)

Fig. 2

Continued fraction solutions: (a) real partand (b) imaginary part for mode j = 1,000.

(b)Fig. 3

Continued fraction solutions: (a) real partand (b) imaginary part for mode j = 0.

For the 2nd soil layer, the higher mode j of 50,000
with λ j = 100,001(π/2) and the very low mode j of 5with λ j = 11(π/2) are chosen for the test. The resultsof continued fraction solutions normalized by λ j forsuch modes are plotted in Figs. 4 and 5, respectively.
Again, at the higher mode, both continued fraction
solutions with the same number of fractions yield
the same accurate result as shown in Fig. 4.
However, at the very low mode, the high-frequency
continued fraction solution with M H = 3 yields thevery poor result as shown in Fig. 5. As the order M H
increases from 3 to 51 and 51 to 113, the accuracy ofthe solution is continuously improved. By contrast,
the doubly asymptotic continued fraction solution
with merely M H = M L = 1 is enough to gain theaccurate result. It is noted that in the case of the
deeper and less permeable soil layer, the order M H
used for the high-frequency continued fraction
solution is much higher when tested at the very low
mode.

Frequency Domain Analysis

(a)

(a)

(a)

(b)

Fig. 4

Continued fraction solutions: (a) real partand (b) imaginary part for mode j = 50,000.

(a)

(b)

Fig. 5

Continued fraction solutions: (a) real partand (b) imaginary part for mode j = 5.

Time Domain Analysis
The accuracy of the modal responses of porewater pressure x� j obtained from the doubly
88GEOMATE- Bangkok, Nov. 14-16, 2016

asymptotic open boundary condition in Eq. (58) isevaluated and compared with that of the highfrequency open boundary condition proposed in [10].
The convolution integral in Eq. (18) as the exact
response is used for verifying the responses. The
small time increment ∆t of 0.005 sec. is used in theanalysis to ensure the accuracy of the results.
Modal seepage flow as double impulse function
In the 1st example, the 1st soil layer is tested at
the high mode j of 1,000 with λ j = 2,001(π/2) andthe lowest mode j of 0 with λ j = π/2 under a shorttime q� j which is the double impulse function asprescribed in Fig. 6(a) with its Fourier transform in
Fig. 6(b). At the high mode, both doubly asymptotic
and high-frequency open boundary conditions using
the same number of fractions yield the sameaccurate response as their curves match the curve of
the exact response in Fig. 7(a). At the lowest mode,the high-frequency open boundary condition with
M H = 3 does not provide an accurate result (see Fig.
7(b)). Only M H of 19 is enough to gain the accuracy.
On the contrary, the doubly asymptotic open
boundary condition with merely M H = M L = 1 still
provides the accurate result.

(a)Fig. 6(a)Fig. 8(b)Bessel function of 1st kind: (a) time historyand (b) Fourier transform.

Both open boundary conditions still provide thevery good results at the high mode as shown in Fig.
9(a). At the lowest mode, the doubly asymptotic
open boundary condition with M H = M L = 1
performs better as its response matches the exact
response while the high-frequency open boundary
condition with M H = 3 yields the poorer result ascompared in Fig. 9(b). Only M H of 19 can match theexact response.

(b)
Double impulse function: (a) time historyand (b) Fourier transform.

(a)

(b)

Fig. 9Modal responses of pore water pressure to
Bessel function of 1st kind: (a) for mode j =1,000 and (b) for mode j = 0.

Modal seepage flow as single impulse function

(a)

(b)

Fig. 7

Modal responses of pore water pressure to
double impulse function: (a) for mode j =1,000 and (b) for mode j = 0.

In the 3rd example, the 2nd soil layer is tested at
the higher mode j of 50,000 with λ j = 100,001(π/2)
and the very low mode j of 5 with λ j = 11(π/2) undera short-time q� j which is the single impulse function
as given in Fig. 10(a) with its Fourier transform in
Fig. 10(b). For the deeper and less permeable soil
layer, both open boundary conditions also yield thesame result at the higher mode as shown in Fig.
11(a). As expected, the doubly asymptotic open
boundary condition with M H = M L = 1 still performs
very well at the very low mode as the response curvefits the exact response as shown in Fig. 11(b). For
the high-frequency open boundary condition, theorder M H of 3 is not enough to gain the accurate
result. It turns out that the order M H must increase to
51 in order to fit the exact curve. This shows thatwhen h increases and c decreases, only the high-Modal seepage flow as Bessel function of 1st kindIn the 2nd example, the 1st soil layer is tested
again at the same modes but under a long-time q� jwhich is the Bessel function of the 1st kind for order� t)̅ where ω� = 30π as shown in Fig. 8(a)1/2 i.e. J1/2 (ωwith its Fourier transform in Fig. 8(b).
89GEOMATE- Bangkok, Nov. 14-16, 2016

frequency open boundary condition is affected
because the order M H used must be higher toguarantee the accuracy of responses.

(a)

(a)

Fig. 13 Modal responses of pore water pressure toearthquake excitation: (a) for mode j =50,000 and (b) for mode j = 5.

(b)Fig. 10 Single impulse function: (a) time historyand (b) Fourier transform.

(a)

CONCLUSIONS
The doubly asymptotic open boundary
condition proposed herein is expressed as a systemof fractional differential equations in the time
domain. It can be solved by the Newmark methodwith the improved numerical method of RiemannLiouville fractional derivative, and the convolution
integral, which is an expensive task, is no longer
required.
In the frequency domain analysis of both soil
layers, the convergence rate of the doubly
asymptotic continued fraction solution is much
faster than that of the high-frequency continued
fraction solution, especially at the low modes. In thetime domain analysis under the short-time and longtime flows, all the results of modal responses of porewater pressure obtained from the doubly asymptotic
open boundary condition using the less number ofcontinued fraction orders are more accurate, in otherwords, the doubly asymptotic open boundary
condition is more efficient than the high-frequency
open boundary condition.

(b)

Fig. 11 Modal responses of pore water pressure to
single impulse function: (a) for mode j =50,000 and (b) for mode j = 5.
Modal seepage flow under earthquake excitation
In the last example, the 2nd soil layer is analyzed
again at the same modes, but using a long-time q� junder earthquake excitation in Fig. 12(a) with its
Fourier transform in Fig. 12(b). For the high mode,
the results obtained from both open boundary
conditions are identical as plotted in Fig. 13(a).
Again, at the very low mode, the high-frequency
open boundary condition is no longer efficientbecause the order M H used is very high, that is, 113
in order to fit the exact curve (see Fig. 13(b)). But inthe case of the doubly asymptotic open boundary
condition, merely M H = M L = 1 is enough to fit theexact curve. This also confirms that for the deeper
and less permeable soil layer, only the highfrequency open boundary condition is affected as theorder M H used must be higher to guarantee theaccuracy of responses.

(a)

(b)

REFERENCES[1] Wolf JP, Song C, Finite-element Modelling of
Unbounded Media. John Wiley & Sons:
Chichester, 1996, ch. 1.
[2] Givoli D, “Non-reflecting boundary conditions:a review”, J. of Computational Physics, Vol. 94,
May 1991, pp. 1-29.
[3] Keller JB, Givoli D, “Exact non-reflecting
boundary conditions”, J. of Computational
Physics, Vol. 82, May 1989, pp. 172-192.
[4] Tsynkov SV, “Artificial boundary conditions
for the numerical simulation of unsteadyacoustic waves”, J. of Computational Physics,
Vol. 189, Aug. 2003, pp. 626-650.
[5] Hagstrom T, De Castro ML, Givoli D, Tzemach
D, “Local high-order absorbing boundary
conditions for time-dependent waves in guides”,
J. of Computational Acoustics, Vol. 15, Mar.
2007, pp. 1-22.
[6] Prempramote S, Song C, Tin-Loi F, Lin G,“High-order doubly asymptotic open boundaries
for scalar wave equation”, Int. J. for Numerical(b)

Fig. 12 Earthquake excitation: (a) time history and(b) Fourier transform.
90

GEOMATE- Bangkok, Nov. 14-16, 2016

Methods in Engineering, Vol. 79, Jul. 2009, pp.
340-374.
[7] Zhao C, Valliappan S, “Transient infinite
elements for seepage problems in infinite
media”, Int. J. for Numerical and Analytical
Methods in Geomechanics, Vol. 17, May 1993,
pp. 323-341.
[8] Jie Y, Jie G, Mao Z, Li G, “Seepage analysis
basedonboundary-fitted
coordinatetransformation method”, Computers andGeotechnics, Vol. 31, Jun. 2004, pp. 279-283.
[9] Chen B, Song E, “An artificial boundary[10] Prempramote

S,“High-frequency

openboundary condition of transient seepage
equation”, in Proc. 19th National Convention on
Civil Engineering, 2014, pp. 1-8.
[11] Zhang W, Shimizu N, “The improved numericalmethod of Riemann-Liouville fractional
derivative for FE formulation”, in Proc. WCCM
VI in conjunction with APCOM’04, 2004, pp.
1-7.
[12] PrempramoteS,“A

high-frequency

openboundary for transient seepage analyses ofsemi-infinite layers by extending the scaled
boundary finite element method”, Int. J. for
Numerical and Analytical Methods inGeomechanics, Vol. 40, Apr. 2016, pp. 919-941.

approach for unbounded transient seepage
problems”, Int. J. for Numerical and Analytical
Methods in Geomechanics, Vol. 39, May 2015,
pp. 762-774.

91

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

3D FINITE ELEMENT ANALYSIS OF EARTH PRESSURE BALANCESHIELD TUNNEL EXCAVATION USING SHELL ELEMENT AND
GROUTING LAYER
P. Lueprasert1, P. Jongpradist2, N. Heama1, K. Ruengwirojjanakul1 and S. Suwansawatt1
Faculty of Engineering, King Mongkut’s Institute of Technology Ladkrabang, Bangkok, Thailand
2
Faculty of Engineering, King Mongkut’s University of Technology Thonburi, Bangkok, Thailand

1

ABSTRACTWith the advancement of computer sciences and researches on tunneling simulation in the past, the 3D finite
element analysis of tunnel excavation by tunnel boring machines (TBMs) has been extensively used over the last
decade. Due to the complicated construction sequences and relevant loads can be taken into account, complex
interaction problems can then be performed. Many simulation techniques have been proposed depending on theassumptions used in the modeling. For modeling the tunnel lining, solid elements are commonly used due to theratio between the width and thickness of the lining is not large. In addition, most studies focused on the grounddeformation, not the lining forces. In the circumstance that the lining forces are essentially observed, thestructural elements that directly provide the values are preferred. Therefore, this research attempts to propose
the shell elements as tunnel lining together with the grouting layer in the analysis. The analysis results from the
proposed method and the conventional one are compared and discussed in terms of ground deformation andlining forces. The analysis focuses on the characteristics of the MRTA tunnel project, (e.g. tunnel diameter andpositions, soil profile and construction process) including the operations of TBMs in the project (e.g. facepressure, shield element or shield weight, tunnel lining and grouting pressure).
Keywords: 3D FEM, Tunnel boring machines, Tunnel lining, Shell element, Lining force
At the beginning, the simulations of tunnelingusing TBMs were analyzed in 2D finite element(FE) models [2]-[5]. In these models, the different
stages of the TBMs advancement are considered bystress-relief method. The 3D FE models for TBMs
tunneling were first applied in the early 1990s.
Various simulation methods considering different
factors and simplifications have been developed [6][10]. Among different factors considered in theanalysis by each researcher, the tunnel lining
element and the tail void grout process have notbeen considered yet. These two parameters were
measured and emphasized by [11] and [12]. The
complicated simulations of 3D EPBS tunnelingmodels using several parameters e.g., face pressure,shield element or shield weight, tunnel lining
element, tail void grout process, hydraulic jack,steering control, and backup trailer loads were
gradually introduced [13], [14]. The 3D FE models
using face pressure, shield element, tunnel lining
element, and grouting process as the parameters areused in studies [15], [16]. As the studies mentioned
above, only four main parameters (face pressure,shield element, grouting process and lining element)
are generally used to model the 3D EPBS tunneling.
Thus, these parameters are used to simulate the 3D
EPBS tunneling in the present work.
However, various methods to model the grouting
process have been proposed in the simulation of the
excavation by EPBS in the past. For example, themethod which uses a distributed pressure to

INTRODUCTION
The construction of tunnels in urban areas can
induce ground movements, which have harmful
effects on existing structures. To limit soil
disturbance and resultant surface settlement, tunnel
boring machines (TBMs) combined with the earthpressure balance shield (EPBS) method are
popularly employed in tunnel construction projects
as shown in Fig. 1. Analyses on ground movement
due to tunneling by TBM or induced structuralforces in the lining are necessary during the design
stage. Among various analysis methods, Finite
Element Method (FEM) is one of the most widelyused tool for tunneling works.

Fig. 1 Schematic section of a Herrenknecht TBM–
EPB machine [1]92GEOMATE- Bangkok, Nov. 14-16, 2016

represent the grouting process was introduced by[17], [18]. In the grouting process in engineering
practice, the grouting will change from the liquidstate to solid. In order to realistically model thischaracteristic, solid elements having different
stiffness are represented to simulate the grouting
process [12], [15] and [19]. However, an actual
behavior of initial grout is in a fluid state with no
stiffness. In practice, it is then difficult to specify themodulus of grouting during liquid state. The
pressure boundary may thus be more appropriate
than the solid elements.
In simulation of ground movement due totunneling, the solid elements or the brick elements
are generally adopted to represent the tunnel lining[13], [14] and [17]. The lining which has a certain
width and the actual excavated periphery of the soil
can then properly be modeled. By using the solid
elements, the structural forces (i.e., bending
moment, normal force and shear force) cannot
directly be obtained. In circumstances that thestructural forces are needed to be investigated, themodeling by the solid elements becomesinconvenient. In this situation, shell elements which
offer the direct quantification of lining structuralforces are preferable. The shell elements have been
also used in the past researches, e.g., the influence ofground stratification on lining force and settlement
due to tunneling [19], the effects of tunneling on
adjacent building and existing tunnel respectively[15]-[16] and [20]. However, using the shellelements to model the tunnel lining causes a
problem due to that the shell elements have nothickness in the process of meshing. It is reasonable
to model the shell element (as line on cross sectional
plane) at the mid plane. Consequently, the treatment
on the gap between the excavated soil periphery andmid plane during meshing is essential.
This study introduces the grouting layer as solid
element in the gap to represent the solid-state
grouting while using pressure boundary to represent
the liquid-state grouting. The comparison of threetunneling simulation methods using 3D FE models
are made and discussed. The first method follows
one suggested in PLAXIS 3D 2013.1 manual [18].
The second method is similar to the first method but
the solid elements as tunnel lining are replaced by
the shell elements assuming the excavated boundary
at the mid plane. The final method, the shellelements used to represent the tunnel lining together
with the adapted grouting process mentioned above.
The results are discussed in terms of surface
settlement profiles and the structural forces in tunnel
liningsimulate in this study as shown in Fig. 2. The
geotechnical conditions for the study site aredetailed by [21]. Geological conditions can be
described as follows:

Fig. 2 Soil profile and pore pressure of case study in
MRTA Blue Line Project
The uppermost first layer consists of weathered
crust or fill material. The second layer is the very
soft clay layer. The first stiff clay is located in thethird layer. A thin seam of clayey sand is foundbelow the first stiff clay as a fourth layer. The
second stiff clay is found below the upper sand.
Tunnel with inner and outer diameter of 5.7 and 6.3,
respectively, is located in the stiff clay layers at thedepth of about 19 m from ground surface. A typicalpore water pressure profile in Bangkok is apiezometric drawdown as shown in Fig. 2. The porewater at the depth of about 20 m is almost zero andrestored condition to hydrostatic pressure at thedepth about lower than 20 m.
FINITE ELEMENT ANALYSIS
Numerical modelFig. 3 The mesh in FE model (the proposed method)

SITE DESCRIPTION

The 3D FE mesh presenting the sample case of
the proposed method (the final method) is depicted
in Fig.3. The soil layers were discretized into the

The section CS-8B of Mass Rapid Transit
Authority (MRTA) Blue Line Project is chosen to93GEOMATE- Bangkok, Nov. 14-16, 2016

six-node bricks or the solid elements with a suitableaspect ratio. The simulation of tunnel components
consisted of three layers, EPB shield layer, grouting
layer and tunnel lining layer. The four-nodes shellelements were used to model the tunnel lining and
EPBS. The hardened grouting layer was simulatedby the solid elements. Their information will be
detailed in next section. The dimension of model is
80 m (12.5DT) in the transverse direction, 60 (9.5DT) in the longitudinal and vertical directions.
The monitoring section is at the center of
longitudinal direction. The distance of 5DT ahead
and behind the monitoring section are enough tofully simulate the tunneling problem. The PLAXIS
3D 2013.1 software was implement to analyze andadopt for mesh generation.

phase, the grout has not yet fully hardened, theliquid state of grouting was simulated by applying aradial pressure with 200 kPa [21] acting on soil
around tunnel. The simulation of tunneling process
in steps 1 and 2 follows ones recommended bymanual of PLAXIS 2013.1 [18]. Step 3, the tail void
grouting in a second phase is considered to behardened, the grouting layer was simulated by thesolid elements. Step 4, the shell or solid elements
representing the tunnel lining was activated in thesame grouting layer section. These steps, 3 and 4,
are differed for the simulation by each method inthis study. The details are described in next section.
The patterns of simulation method

Earth pressure balance shield advancement andsimulation procedures
The tunneling process of EPBS was simulatedusing a step-by-step approach. Each excavation stepcorresponded to an advancement of the tunnel faceof 1.2 m which is equal to the width of tunnel lining.
A simplified geometry of EPBS is assumed with theoriginal cone-shaped replaced by a cylindrical shape.
The schematic of simulated process with EPBS wasshown in Fig. 4. The simulation process can be
described as follows.

Fig. 4 The tunneling simulation process
Step 1, the soil elements were deactivated as the
EPBS diameter which the over cutting of shield did
not simulate in this study. The support of excavation
face was modeled by applying a pressure
distribution with linear increase of pressure withdepth. The face pressure in this study is about 150 to
200 kPa at crown and invert of tunnel respectively[21]. The shell elements were activated to represent
the EPB shield with contraction ratio of 0.4%, which
was calibrated from the previous FE analysis oftunneling projects in Bangkok subsoil. These
procedures were repeated until the advancement ofshield was completed with seventh rings for thewidth of about 8.4 m in longitudinal direction. Step
2, the simulation of tail void grouting in a first

Fig. 5 The cross sections of simulation patterns
The simulation method with EPBS tunneling is
divided into three patterns. The difference ofpatterns is that the different techniques are selected
to simulate the tunnel lining and grouting layer
section as shown in Fig. 5.
The first method that is called “NG-SOLID
method”, the tunnel linings are modeled by the solid
elements with thickness of 0.30 m. The step 3 ofsimulation procedures is not considered in thismethod. The second method, that is called “NGSHELL method”, is similar to the first method. The
shell elements are represented to simulate the tunnel
94GEOMATE- Bangkok, Nov. 14-16, 2016

linings, the geometry of simulation is thus different.
The difference between using the solid or shellelements is that the geometry of thin shell element ismodeled as zero-thick line at mid plane. Thus, thediameter of excavated soil periphery in thismodeling is 6.0 m with the same as shield diameter.
The details of two simulation methods are depicted
in Fig 5.
The final method was suggested in this study.
The shell elements are used to represent the tunnel
lining together with the grouting layer, that is called“AG-SHELL method”. The thickness of grouting
layer is 0.07 m which is equal with the thickness of
EPB shield in the case that the pitching angle ofexcavation process and over cutting of EPBS is notconsidered in this study. The cross section of AGSHELL method is depicted in Fig. 5b.

The properties of soil are determined from the
MRTA projects [22]. The soil layers, stiff and soft
clays, were assumed by hardening soil model (HS)[23]. The Mohr-Coulomb model (MC) was assumed
to represent the weathered clay and sand layer. The
soil properties in this study are calibrated material
parameters for the accuracy of simulations ofgeotechnical work in Bangkok subsoil [24] as shownin Table 1.
Table 2 Material properties of EPB shield, tunnel
lining and grouting layer.
EPB
Elements

Analysis condition

Young
modulus[E](kN/m2)Unit
weight[ ]Tunnel lining

31 x 1060.20(kN/m3)24

EPB shield

210 x 106

0.28

78

1 x 106

0.30

21Grouting layer

The initial distribution of vertical effective stressand horizontal effective stress are controlled by thegiven soil unit weight, the coefficient of earthpressure at rest, K0 for all strata and the hydrostatic
pore water pressure conditions in equilibrium with awater table at the ground surface. The undrained
analysis was considered.
The displacement boundary was adopted in thisstudy. The sides of the mesh including the front sideand rear side are restrained against lateral
movements but free to move vertically, so no
movement perpendicular to their side of meshes.
The bottom of the mesh is fixed (no vertical andhorizontal movements). These conditions were used
for all finite element meshes throughout of theanalysis.

Poisson’s
ratio
[ ]

Table 2 shows the properties of the components
of EPBS tunneling simulation. The EPB shield,
tunnel lining and grouting layer were assumed to belinear elastic. The properties of grouting layer at 28
days of curing are obtained from [13], [14] while
those of EPBS are acquired from [19].
ANALYSIS RESULTThe results in terms of surface settlement andlining forces from the three simulation methods arecompared and discussed together with themeasurement data if available. The observations ofFE analysis results are carried out after the processof tunneling simulation is completed.

Material properties
Surface settlement
Table 1 Soil parameters for modeling [22]Soil layer

Wea.
crust

Material
model

MC

Soft
clay

Stiff
clay

Hardening Soils(HS) model
-

Sand

MC

E ' (kPa)
ref
(kPa)
Eoed

6,000
-

5,000

60,000-ref
50 (kPa)-

5,000

60,000-refur (kPa)-

15,000180,000

-

17

16

18

20

0.32

0.33

0.33

0.3

o

22

22

22

36

c (kPa)
m (-)

8
-

5
1

18
1

0
-

pref (kPa)

-

100

95

-

E
E

 sat (kN/m ) ' (-)
' ( )
3

80,000Fig. 6 Comparing transverse settlement profile from
MTRA section CS-8B with FEM analyses
Figure 6 shows the surface settlement profiles
compared between FE analyses and MRTA
monitoring data in section CS-8B. The surface
settlement profiles analyzed by FEM are similar inshape but quantitatively different. The surface
settlement profile of the NG-SOLID method isnoticeably deeper than the surface settlement95

GEOMATE- Bangkok, Nov. 14-16, 2016

obtained from the NG-SHELL method. It is clear,
the selection of simulation method significantly
affects on the nature of the computed surface
settlements. The difference of the surface settlement
profiles may be presumed to be associated withproperties of the elements or the excavated cavity oftunneling.
Although the excavated cavity of NG-SOLID islarger than the AG-SHELL method, the surface
settlement profile of the AG-SHELL method is close
to that of the NG-SOLID and the MRTA monitoringdata. This may attribute to the grouting layer (solid
element) adapted in the AG-SHELL method.

Axial force
Figure 8 depicts the computed axial force
resulted by three simulation methods. The tendencyof the computed axial force is similar to thecomputed bending moment. The computed axial
force of the NG-SOLID and the AG-SHELL
methods is in the same range. This range agrees wellwith the previous researches [25], [26], whichshould be in the range of 0 to 1000 kN/m. In contrast,
the distribution of the computed axial force of the
NG-SHELL method is in much higher.

Lining force
The comparison of the computed structural forcesin the tunnel lining for three methods is presented
hereafter. The structural forces are plotted from thereference lining ring located at the mid-section ofmodel.
Bending moment

Fig. 8 The computed axial force in linings for threedifferent methods
CONCLUSION
In this study, we proposed to model the tunnel
lining with the shell elements together with the
grouting layer in tunneling analysis by TBM in soft
ground. The grouting is modeled as pressure
boundary and solid elements in liquid and solidstate, respectively.
By the proposed method,
grouting process and actual excavated boundary of
the soil can then be realistically modeled. Analysisof tunnel excavation using the proposed method in
conjunction to 4 main factors (face pressure, shieldelement, grouting process and lining element),
provides good agreement with results from theconventional method. Besides, the structural forces
obtained from the current method are in ranges ofhistory records.

Fig. 7 The computed bending moment in linings forthree different methods
Figure 7 shows the computed bending moment
resulted by the three simulation methods. From thisfigure, the significant differences of computed
bending moment are revealed. Especially, thedistribution of the computed bending moment by the
NG-SHELL method is drastically different withthose by the others. The computed bending momentof tunneling simulation located in soft clay (K0 <1.0) is positive values at the spring line and negativevalues at crown and invert. This behavior obtainedby the computed bending moment of NG-SOLID
and AG-SHELL methods. Although the trends of thecomputed bending moment for NG-SOLID and AGSHELL methods are similar, the magnitudes aredifferent. However, the ranges of magnitude ofcomputed bending moment obtained from the NGSOLID and AG-SHELL methods are close to thosereported in the previous studies [25], [26].

ACKNOWLEDGEMENT
The financial support provided by the researchfunding from the Thailand Research Fund (TRF) and
1D2 Group Co. Ltd. under Grant PHD56I0057 isacknowledged.
REFFERENCES[1] Herrenknecht AG, “Herrenknecht” [Online],
Germany,Available:https://www.herrenknecht.com/en/products/core-products/tunnelling/epbshield.html [5 April 2016].

96

GEOMATE- Bangkok, Nov. 14-16, 2016[16] Zheng G, Zhang T, Diao Y, “Mechanism andcountermeasures of preceding tunnel distortion
induced by succeeding EPBS tunnelling in close
proximity”, Computers and Geotechnics, 66,
2015, pp. 53-65.
[17] Broere, W, Brinkgreve, R.B.J., ‘‘Phased
simulation of a tunnel boring process in soft
soil’’, in Proc. Conf. on Numerical Methods inGeotechnical Engineering, in Mestat (Ed.),
Presses de 1’ENPC/ LCPC, Paris, 2002, pp.529536.
[18] PLAXIS Version 8 Manual. A.A. Balkema
Publishers, 2013[19] Katebi H, Rezaei A, Hajialilue-Bonab M,
Tarifard A, “Assessment the influence ofground stratification, tunnel and surfacebuildings specifications on shield tunnel lining
loads (by FEM)”, Tunnelling and UndergroundSpace Technology, 49, 2015, pp. 67-78.
[20] Comodromos
EM,
Papadopoulou
MC,
Konstantinidis GK, “Numerical assessment ofsubsidence and adjacent building movements
induced by TBM-EPB tunneling”, J. of
GeotechnicalandGeoenvironmental
Engineering, 140 (11), 2014[21] Suwansawat. S. “Earth Pressure Balance (EPB)
Shield Tunneling in Bangkok: GroundResponse and Prediction of Surface Settlements
Using Artificial Neural Networks”, Ph.D.
Thesis, Massachusetts Institute of Technology,
USA, 2002.
[22] Jongpradist P, Kaewsri T, Sawatparnich A,
Suwansawat S, Youwai S, Kongkitkul W,
Sunitsakul J, “Development of tunnelinginfluence zones for adjacent pile foundations by
numerical
analyses”,
TunnellingandUnderground Space Technology, 34, 2013, pp.
96-109[23] Schanz T, Vermeer PA, Bonnier PG, “Thehardening soil model: formulation andverification”, in Proc. Conf. on Beyond 2000 in
Computational Geotechnics-10 years of Plaxis,
Balkema, Rotterdam, 1999, pp. 281–296.
[24] Prust RE, Davies J, Hu S, “Pressure meter[2] Clough GW, Sweeney BP, Finno RJ,“Measured soil response to EPB shieldtunneling”, J. of Geotechnical EngineeringASCE, 109(2), 1983, pp. 131-149.
[3] Finno RJ, Clough GW, “Evaluation of soil
response to EPB shield tunneling”, J. of
Geotechnical Engineering-ASCE, 111(2), 1985,
pp. 155-173.
[4] Bernat S, Cambou B, “Soil-structure interaction
in shield tunnelling in soft soil”, Computers andGeotechnics, 22(3/4), 1998, pp. 221-242.
[5] Bernat S, Cambou B, Dubois P, “Assessing asoft soil tunnelling numerical model using fielddata”, Geotechnique, 49, 1999, pp.427-452.
[6] Lee KM, Rowe RK, “Finite element modellingof the three-dimensional ground deformations
due to tunnelling in soft cohesive soils: part I –method of analysis”, Computers andGeotechnics, 10, 1990, pp.87–109.
[7] Lee KM, Rowe RK, “Finite element modellingof the three-dimensional ground deformations
due to tunnelling in soft cohesive soils: part II results”, Computers and Geotechnics, 10, 1990,
pp.111-138.
[8] Lee KM, Rowe RK, “An analysis of threedimensional ground movements: The Thunder
Bay tunnel”, Canadian Geotechnical Journal, 28,
1991, pp.25-41.
[9] Lee KM, Rowe RK, Lo KY, “Subsidence owing
to tunnelling. I. Estimating the gap parameter”,
Canadian Geotechnical Journal, 29, 1992,
pp.929-940.
[10] Rowe RK, Lee KM, “Subsidence owing totunnelling. II. Evaluation of a prediction
technique”, Canadian Geotechnical Journal, 29,
1992, pp.941-954.
[11] Abu-Krisha A. Numerical modelling of TBM
tunnelling in consolidated clay. Ph.D. Thesis,
University of Innsbruck, 1998.
[12] Swoboda G, Abu-Krisha A, “Three-dimensional
numerical modelling for TBM tunnelling inconsolidated clay”, Tunnelling and UndergroundSpace Technology, 14(3), 1999, pp. 327-333.
[13] Kasper T, Meschke G, “A 3D finite element
simulation model for TBM tunnelling in soft
ground”, J. Int. for Numerical and Analytical
Methods in Geomechanics, 28, 2004, pp.14411460.
[14] Kasper T, Meschke G, “A numerical study ofthe effect of soil and grout material properties
and cover depth in shield tunneling” Computers
and Geotechnics, 33, 2006, pp. 234-247.
[15] Chakeri H, Ozcelik Y, Unver B, “Effects ofimportant factors on surface settlement
prediction for metro tunnel excavated by EPB”,
Tunnelling and Underground Space Technology,
36, 2013, pp. 14-23.

investigation for mass rapid transit in Bangkok,
Thailand”, J. of the transportation researchboard, Transportation research of the national
academies, Washington, DC, 1928, 2005, pp.
207–217.
[25] Working Group No. 2, ITA, “Guidelines for the
Design of Shield Tunnel Lining”, Elsevier
Science Ltd: Published, 2000
[26] Moller S, “Tunnel induced settlements andstructural forces in linings” Doctoral Thesis,
University of Stuttgart, Germany, 200697

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ANALYSIS OF TUNNELING CONSTRUCTION WITH GROUNDSPRING MODEL
1

P. Chaipanna 1, P. Jongpradist 1 and M. Sugimoto 2
King Mongkut’s University of Technology Thonburi, Bangkok, Thailand
2
Nagaoka University of Technology, Niigata, Japan

ABSTRACTThe shield tunnel has been extensively implemented for underground transportation of urban areas. Many
reports of tunnel segment damage during tunnel construction. The jack thrust force is the essential causes ofdamage to the segment. The objective of this article is to present the model which analyzes the behavior oftunnel lining during construction. This model consists of five rings and ten segments for each ring. The segments
are connected with the shear springs which represent the segment joint and ring joint. The zero thickness
interface elements assigned is used to model the segments interface behavior. The jack trust force is applied at
the first ring. The lubricant pressure, grout pressure and soil pressure are taken into account in this model. The
ground spring model is implemented into FEM program ABAQUS, then the soil stiffness is represented by meanof spring. This analysis is separated into two cases; straight alignment and curve alignment. From the analysis
result, the analysis result agrees with the measurement, the stresses of tunnel segment are governed by jack thrustforce. The ground-spring model has been proven to be appropriate for analysis of suitable behavior of the tunnel
segment.
Keywords: 3D FEM, Soil-Tunnel Interaction, Ground spring model, and Tunneling construction

rational to represent the natural soil.
Although, it is commonly realized that the effectof the tunnel construction loads on the tunnel lining
must be taken into account in the analysis and designof shield tunnels, only few research has been carriedout to grasp the effect of tunnel construction loads [5,6, 8]. This paper presents a three-dimensional
analysis model to enhance the appreciation of tunnel
lining behavior due to the construction loads. The
measurement data were taken from the literature of
Nishin-Shinjuku [9] tunnel allows the verification of
the present analysis.

INTRODUCTION
The shield tunneling technology is continuously
developed, the shield tunneling is possibly
constructed in the difficult ground conditions atincreasing soil and water pressures, and various
cross-section of the tunnel. The segmental lining is
popularly used in the shield tunneling. The segments
are sequentially erected as tunnel boring advances,which are consecutively coupled with the bolts inthe longitudinal and circumferential direction of thetunnel. The packing materials are used between
segments with the purpose of load distribution.
During the shield tunnel excavation, the tunnel
linings are subjected to the construction loads, suchas, the jack thrust, the shield tail pressure, thegrouting pressure and ground pressure. Many reportsof shield segment damage during construction havebeen increasingly presented and these indicate thesignificance of construction load [1, 2]. However,the conventional design of tunnel lining takes into
account only the serviceability loads in shieldtunneling [3, 4].
Nowadays, analysis of shield tunneling has been
already achieved by mean of three-dimensional (3D)
model [5,6]. These analyses take into account thestep of tunnel construction, but the tunnel lining ismodeled by means of continuous element. Katebi etal.,[7] and Blom et al.,[8] model tunnel lining bysegmental lining, but the surrounding soil ischaracterized by elastic behavior which is not

FIELD MEASUREMENT
In order to verify the appropriateness of thepresent analysis, the field measured data during theconstruction of Nishin-Shinjuku tunnel have beenused to validate the proposed model [9]. The NishinShinjuku tunnel is a part of the Metropolitan
Expressway Central Ring Shinjuku Line. The
Nishin-Shinjuku Tunnel comprises of twin tunnels
and a total length of 600 m. The outer diameter oftunnel is 13.0 m, thickness is 0.55 m and segmental
width is 1.2 m. For field instrumentation, strain
gauges were installed in first three tunnel rings, 10
positions for each ring.
The tunnel crown is located at the depth of 22 m
while groundwater table is at 13 m below the groundlevel. The tunnel was excavated through very dense
gravel and very dense sand. The geotechnical data98

GEOMATE- Bangkok, Nov. 14-16, 2016

are tabulated in Table 1.
Table 1 Soil properties of analysis [9]

50

Ring
JointsANALYSIS STRATEGYThis section presents a three-dimensional
analysis for shield tunneling, which takes into
account the soil and water pressure, the tail void
grouting pressure, the wire brush grease pressure,the jack thrust and soil-tunnel interaction. The
analysis is performed by means of FE analysis.
The tunnel lining considered in the analysis
includes five rings, each ring comprises of ten
segments, which are staggered alignment along thetunnel axis. The segments are coupled with thesegment joints and ring joints in longitudinal andcircumferential directions, respectively. These joints
are considered using the shear spring elements. The
assembly of tunnel lining is illustrated in Fig.1(a).
The attribution of packer material is represented
with the interface element which involves thenormal and tangential behaviors.
There are various loads acting on the tunnel
lining, such as, jack thrust, wire brush pressure, tailvoid grouting pressure and earth pressure. These
loads influence on the performance of the tunnel
lining. For this reason, these loads are taken into
account in the analysis, as seen Fig.1(b).
The soil-tunnel interaction is modeled by meansof a series of radial spring [10], the stiffness of radialspring is regarded using nonlinear elastic behavior(Fig. 2). The stiffness of radial springs are varied
along the longitudinal tunnel, as seen in Fig.1(c).
The Ring1 in the model is only subjected by thejack thrust in longitudinal direction, no any loads
acting on radial direction. The grease is injected into
the gap between shield tail and the extrados oftunnel lining to prevent the external water flow into
the shield. The grease pressure acts on the Ring2.
Ring3 is subjected by tail void grouting pressure.
The grout material is filled the gap between theextrados of tunnel lining and surrounding soil. The
constant pressure is kept over the hydrostaticpressure of 90 kPa. Ring4 and Ring5 are subjectedto the surrounding ground pressure which comprises
the soil pressure and water pressure.
The simulation of shield tunneling is divided in
two categories, the shield tunneling for straightforward alignment and the curve forward alignment.
The hydraulic jack pattern for both categories areillustrated in Fig.3. The black circles are represented
the active jack thrust, in the other hand, the whitecircles are represented the inactive jack thrust.

Ring3
Ring4

Wire
brush
Grouting(b) Loads distribution

Ring2

Earth
pressure
Ring4

Ring1

Grouting
Ring3

Ring5

Grease
Ring2

Ring1

Jack Thrust(a) Configuration of ring and segment joints
GroundspringN value

Value
18.6
38.0
37.0
91.0

Ring5

Earth layer
Wet density (kN/m3)
Adhesive force (kN/m2)
Internal friction angle (deg)Subgrade reaction coefficient (MN/m3)

Segmental
Joints(c) Radial springs

Fig. 1 Loads distribution and boundary conditionof analysis model
Radial force, fr (kN)

0 Deg.
24 Deg.
48 Deg.
72 Deg.
96 Deg.

40

30

20

10

0
0.000

-0.005

0.005

0.010

0.015

0.020

0.025

Radial displacement, δr (m)-10Fig. 2 Modeling of the soil-tunnel interaction
1

1
7
35

35

Total Trust
94540 kN

Total Trust
83370 kN

14

14
28

28
21

(a) Straight alignment
Fig. 39921(b) Curve alignment

Configuration of jack thrust.

GEOMATE- Bangkok, Nov. 14-16, 2016

from curve alignment (Fig. 7(b)).
The distribution of the circumferential stress isshown in Figs.8 (a)-(b) for straight and curvealignments, respectively. The jack thrust has strong
influence on the circumferential stress. The
circumferential stress mostly concentrated on the

ANALYSIS RESULTThe simulation results of Ring1, Ring2, andRing3 are compared to the field measurements takenfrom literature to validate and calibrate this analysis
method.
The measured and analysis results from Figs.4 to
7 are traced from centerline of tunnel segments. The
measured and computed results are compared bymean of plotting the results (bending moment, forcesand deformation) in the circumferential direction
against the circumferential angle. The contour
shedding of stress distribution are also depicted as
illustrated in Figs.8 to 10.
Fig. 4 shows the comparison between themeasured and computed of circumferential bending
moment of the tunnel Ring1, Ring2 and Ring3. The
left figures represent results of simulation during
straight alignment, and the right figure is for shieldtunneling
duringcurvealignment.
The
circumferential bending moment occurs slightly in
the Ring1 because there is no external force incircumferential direction; only gravitational force ofthe segments themselves exists. The jact thrusts arein longitudinal direction. The moment becomesincreasing in Ring 2 and Ring 3due to the grease andgrouting pressures, respectively. Comparison
between the computed and measured results, it isseen that the analysis can reproduce the moment
distribution in a fair agreement with te measured
ones. The difference between the straight and curvealignment can be also reflected.
Fig. 5 summarizes the computed circumferential
bending from all 5 rings. The maximum
circumferential bending occur in the Ring 5 as aresult of different vertical and horizontal pressure ofsurrounding soil. The influence of jack thrust (asstraight or curve alignment) on the bending is clearlyseen as shown in Figs. 5(a) – (b).
Fig. 6 shows distribution of circumeferential
force comparing between the measured andcomputed values. The maximum circumferential
force appears in Ring2. For Ring 1 and Ring 2, it isseen that the computed forces are in good agreementwith the measured one, whereas, the computed
forces in Ring 3 are larger than those of themeasurement.
From figs.4 and 6, it indicates that the analysis
results satisfactorily correspond to the measurement
for both bending moment and circumferential force.
Fig. 7 depicts the computed circumferential
forces from all 5 rings. The maximum
circumferential load occurs at the invert of Ring5.
The jack thrust does not only affect the axial load,
but also the the circumferential load. Furthermore, acircumferential load of Ring5 is affected from thejack thrust too. The distribution of circumferential of
Ring5 of straight alignment (Fig. 7(a)) differ slightly

200

200

0

0

-200

-200

-400

-400

-400

-400

-200

-200

0

0

200

200

(a) Ring1

Field Measurement
Present Analysis

200

200

0

0

-200

-200

-400

-400

-400

-400

-200

-200

0

0

200

200

(b) Ring2

Field Measurement
Present Analysis

200

200

0

0

-200

-200

-400

-400

-400

-400

-200

-200

0

0

200

200

(c) Ring3

Field Measurement
Present Analysis

Fig.4
Comparison of bending moment (kN-m)
between field measurement and present analysis(Left side for straight alignment and right for curvealignment).
200

200

0

0

-200

-200

-400

-400

-200

-200

0

0

200

200

Ring1,
Ring4,(a) Straight alignment

Ring2,
Ring5

Ring3,(b) Curve alignment

Fig. 5 Computed bending moment distribution(kN-m).
100

GEOMATE- Bangkok, Nov. 14-16, 2016alignment).
front face of Ring1 which is in accordance with theposition of hydraulic jacks. The circumferential
stress decreases in Ring2 and Ring3. The
circumferential stress increases again in Ring4 andRing5 due to radial pressure. The eccentric jackthrust obviously shows the difference of stressintensity between the straight and curve alignment.
The jack thrust directly affects to the longitudinal
stress, this stress concentrates at the hydraulic jackposition on the front face of Ring1. The longitudinal
stress gradually decreases in the other rings. The
contour shading of longitudinal stress of straight andcurve alignments obviously differ between thestraight and curve alignment of the analysis, see
Figs.9(a)-(b).
The computed shear stress of tunnel lining isshown in Figs.10(a)-(b). Due to the tunnel lining issegmental, the shear stress of adjacent segments aretransferred by mean of interface elements and tunnel
joint. The maximum shear stress occurs at theposition of hydraulic jack.
2000

2000

1000

1000

0

0

-1000

-1000

-2000

-2000

-2000

-2000

-1000

-1000

0

0

1000

1000

4000

4000

2000

2000

0

0

-2000

-2000

0

0

2000

2000

4000

4000

Ring1,
Ring4,(a) Straight alignment
Fig. 7Ring3,(b) Curve alignment

Computed circumferential force (kN).

a) Straight alignment
Fig. 8Ring2,Ring5b) Curve alignment

Distribution of circumferential Stress (kPa).

2000

2000

(a) Ring1

Field Measurement
Present Analysis

2000

2000

1000

1000

0

0

-1000

-1000

-2000

-2000

-2000

-2000

-1000

-1000

0

0

1000

1000

a) Straight alignment
Fig. 9b) Curve alignment

Distribution of longitudinal stress (kPa).

2000

2000

(b) Ring2

Field Measurement
Present Analysis

2000

2000
0

0

-2000

-2000

-4000

-4000

-4000

-4000

-2000

-2000

0

0

2000

2000

(c) Ring3a) Straight alignmentb) Curve alignment

Fig. 10 Distribution of shear stress (kPa).
Fig. 11, the differential pressure acts on theextrados of tunnel rings that heavily affect to its
radial displacement. The Ring5 shows maximum
radial displacement that is induced by the groundpressure. The diameter of tunnel lining increases in
the horizontal direction and decreases in the verticaldirection. The overall tunnel diameter reduces due toField Measurement
Present Analysis

Fig. 6 Comparison of circumferential force (kN)
between field measurement and present analysis(Left side for straight alignment and right for curve101GEOMATE- Bangkok, Nov. 14-16, 2016surrounding pressure.

REFERENCES

0.004

0.004

0.000

0.000

-0.004-0.004-0.008-0.008-0.004-0.004

0.000

0.000

0.004

[1] Sugimoto, M., “Causes of Shield Segment
Damages During Construction”, in Proc. Int.
Symp. On Underground Excavation and
Tunnelling, Bangkok, 2006,Thai, pp.67-74[2] Wu M.Q., “Application study on steel fiber
concrete segment in metro tunnel engineering”,
Guangdon Building Materials 3, 2004
[3] ITA-WG2, “Guidelines for the Design of Shield
Tunnel Lining”, Tunnelling and UndergroundSpace Technology, Vol.15, No.3, 2000, pp.303331[4] Japanese Society of Civil Engineerings (JSCE),“Japanese Standard for Shield Tunneling”, 3rded. 1996, Tokyo, Japan[5] Kasper T. and Meschke G., “A 3D finite
element simulation model for TBM tunnellingin soft groung”, Int. J. Numer. Anal. Meth.
Geomech. 28, 2004
[6] Katebi H., Rezaei A.H., Hajialilue-Bonab M.
and Tarifard A., “Assessment the influence ofground stratification, tunnel and surfacebuildings specifications on shield tunnel lining
loads (by FEM)”, Int. J. Tunn. Undergr. SpaceTechno. 49, 2015[7] Blom C.B.M., van der Horst E.J. and Jovanovic
P.S., “Three-dimensional Structural Analyses of
the Shield-Driven Green Heart Tunnel of theHigh-Speed Line South”, Int. J. Tunn. Undergr.
Space Techno. 12, 1999
[8] Mo H.H. and Chen J.S., ”Study on inner force
and dislocation of segments caused by shieldmachine attitude”, Int. J. Tunn. Undergr. SpaceTechno. 23, 2008
[9] Tajima M., Kishida M., Fukai N. and Saitou M.,“Study on shield tunnel construction loads usinga three-dimensional FEM model”, in Proc.
Tunnel Engineering, JSCE Vol.14, 2004,
pp.353-360[10] Chaipanna P., Jonpradist P. and Kalasin T., “3D
FEA of Shield Segmental Lining with Ground
Spring Model”, in Proc. ITA-AITES World
Tunnel Congress 2012

0.004

Ring1,
Ring4,(a) Straight alignment

Ring2,
Ring5

Ring3,(b) Curve alignment

Fig. 11 Computed radial displacement of theanalysis
CONCLUSIONS
A three-dimensional analysis of shield tunnel
segments during construction has been presented,
which takes into account all relevant interactions andloads, i.e., soil and water pressures, the grouting
pressure, the jack thrust force and the greasepressure. The conclusions from this study are as
follows:- The
present
analysis
satisfactorily
correspond to the field measurement of casestudy.
- The analysis results can reproduce theinfluence of jack thrust on the tunnelingbehavior.
- The jack thrust influence on both
displacement and structural forces of tunnel
lining.
- The jack thrust is the main cause of the
maximum stress in the segment. Therefore,
the internal stress of tunnel lining isgoverned by the jack thrust.

102

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THERMO-MECHANICAL ANALYSIS OF GEOTHERMAL PILES IN
DENSE SAND
1,2

Anant Aishwarya Dubey1, Suresh Kumar2
Department of Civil Engineering, Indian Institute of Technology (BHU), Varanasi, India.

ABSTRACT
Increasing population and leading development has impacted the surrounding environment with multiplied
pollution in the attempt to meet the energy demands. To reduce the harm causing pollutant, the advancement ofsustainable and renewable energy sources, is essential. One of the most reliable source of renewable energy is
geothermal energy and with geothermal structures, it is possible to transfer energy from the ground to fluid-filledpipes casted in concrete and then to building environments. For efficient design of geothermal piles, it isnecessary to evaluate the thermo-mechanical behavior of geothermal piles in local conditions. The aim of thisstudy is to evaluate the behavior of geothermal pile subjected to thermo-mechanical loading. First, in this study
the field results are validated and then parametric study by varying mechanical and thermal loading on thegeothermal pile placed in dense sand, is performed. In this study the sand-geothermal pile is modeled as axissymmetrical model where pile is considered to be thermo-elastic in nature and sand is modeled as MohrCoulomb elastic-plastic material. Coupled Temperature-Displacement step is applied in order to simulate theobserved experimental results. It is shown that the numerical model is able to reproduce the major thermomechanical effects approximately.
Keywords: Geothermal Piles, Numerical Modelling, Thermo-Mechanical Analysis, thermal response of pile.
intermittently at different times of the year over
several years’ operation of the GSHP system.
Second field trial on geothermal piles wasundertaken at E´cole Polytechnique Fe´de´rale deLausanne (EPFL) by Laloui et al. (2003, 2006),
Switzerland, during the construction of a newbuilding on the campus. They reported pile uplift
and temperature-induced axial stress in the pile ishigher as compared to that caused by the mechanicalload.
Later Amis et al. (2008) and Bourne-Webb et al.
(2009) performed energy pile field trials in Lambeth
College, London, UK. The test site is located within
the grounds of the Clapham Centre of Lambeth
College in South London.
McCartney et al. (2010) performed centrifuge model
tests of thermo-active foundations under 60 °C
heating and subsequent cooling to 25 °C andobserved an increase in shaft capacity of the piles
due to heating. Wang et al. (2011) analysed theeffect of coupled thermo-mechanical loading on theload carrying capacity of laboratory scale model
geothermal energy piles. They reported thattemperature is an important parameter which
influences the load carrying capacity of the piles.
McCartney and Murphy (2012) performedexperiments on two full-scale energy foundations
installed at the new Denver Housing Authority
Senior Living Facility in Denver, Colorado. Upward
expansion of the piles during heating was observedfor end-bearing piles. Centrifuge test on energy piles
in sand has been performed by Goode et al. (2014).
They observed upward displacement of the pile head

INTRODUCTION
The geothermal piles function as sub-structural
support to the building and also fulfill the building’s
heating–cooling requirements. For manufacturing ofenergy piles, the reinforced concrete drilled shafts
are often preferred over the steel pipe piles becauseof good thermal storage capacity and thermal
conductivity of concrete that make these piles anideal medium for geothermal energy absorber(Brandl 2006). In spite of the growing
acknowledgement on the benefits of energy piles,
there is still knowledge cavity in understanding theload transfer mechanism of these piles underthermo-mechanical loading (Laloui et al, 2006).
There are three major ways to analyse the behaviourof geothermal piles, first one is full scale field in-situ
tests; the second is, Laboratory scale modelling tests
and Centrifuge Modelling tests; and the third way is
Theoretical analysis and Numerical Simulations.
First full scale field in-situ test on thermomechanical behaviour of geothermal pile was
performed at Bad Schallerbach, Austria by Brandl et
Al (1998, 2006). An instrumented pile, forming partof an operational GSHP system of 143 piles, wasinstalled during construction works for arehabilitation centre in Bad Schallerbach, Austria(Brandl, 1998, 2006). The piles are part of a piled
raft, supporting a seven-storey building that isbenched into a sloping site. The test pile was
equipped with pressure cells at the toe and the head,‘fissure meters’ at three levels, and ‘thermoelements’ at five levels. Data were collected
103

GEOMATE- Bangkok, Nov. 14-16, 2016

and uniform thermal axial strain in the pile due toheating of the pile.
Kramer and Basu (2014) at The Pennsylvania State
University performed laboratory scale modelling ofgeothermal pile by applying thermal load on pile via
fluid (ethyl-glycol and water) through a inbuilt PVC
pipe in the energy pile model itself to simulate thereal phenomena of thermal load transfer mechanismof energy piles. They concluded increase in velocity
of circulating fluid increases the thermal
performance of the energy pile, however, sincehigher velocity of fluid consumes more electricity, it
may have unfavorable effect on seasonal
performance factor on the system. They also
reported that variation of circulation speed doesn’t
affect the temperature of surrounding soil
significantly. It was confirmed by the laboratory
scale model testing that the positive and negative
temperature gradient have equal and opposite effects
on the heat transfer and on soil temperature
increment.
Another effective way to analyse the ThermoMechanical behaviour of energy pile is bytheoritically studying the complex behavior ofenergy pile, and/or Numerical Simulation. Professor
Lyesse Laloui, Alice Di Diona, M. E.Suryatriyastuti,
Knellwolf, and Prof. Guney Olgun have performed
the simulation and analysis of Geothermal Piles on
various software packages like Lagamine finite
element software, Ansys, COMSOL and Abaqus
Simulia. They have reported on various aspect ofgeothermal piles like Shaft resistance, Axial Load.
Stress behavior and Thermal Performance of energy
piles. Laloui et al. (2006) performed finite element(FE) analyses of energy piles. They noted that thethermal piles may undergo uplift as opposed to themechanical piles which settles when subjected tostatic dead weight of building. Knellwolf et al.
(2011) proposed a new finite difference analysis
method for geothermal piles considering elastic pile
behavior and neglecting the radial strain in pileduring heating and cooling. Amatya et al. (2012)
analyzed the results obtained from field tests onenergy piles during heating and cooling. Theyobserved that the pile temperature after a thermal
recovery cycle was always slightly higher than theinitial temperature in each test phase, indicating abuild-up of thermal mass around the test pile. The
Ground Source Heat Pump (GSHP) association
proposed design charts (GSHP 2012) for predicting
axial stress and pile head displacement of thermal
piles in London clay using a thermal pile software.
They predicted that the axial stress in the pile andpile uplift increase with increasing pile temperature.
Ghesami-Fare and Basu (2013) performed heattransfer analysis on geothermal energy piles using

finite difference method & reported that heat transferfrom energy piles happens in radial direction.
Mimouni and Laloui (2013) studied the effect ofradial strains in thermal piles on the associated soil–
structure interaction numerically. Arson et al. (2013)
investigated the effect of debonding and presence of
air pocket at the soil– pile interface in heatpropagation from the geother- mal piles in dry sand.
They observed that air pocket at the pile–soilinterface has an insulating effect and thus, less heatis transmitted to the ground than in the case of
perfect adhesion, which results in higher temperaturein the pile and lesser temperature in the soil. They
also observed that debonding has a critical effect on
the mechanical performance of the pile due to loss offrictional resistance. Di Donna and Laloui (2013)
studied the settlement behavior of energy pile due tocyclic heating of the pile and noted that the pilesettlement increases only for the first five cycles ofheating and then settlement becomes constant. They
also reported that thermal load may cause anadditional settlement in the order of 10 % of thesettlement induced due to mechanical load. Wang etal. (2012, 2014) performed coupled thermo-poremechanical FE analysis of centrifuge experiments onenergy piles in saturated and partially saturated silt,
respectively. They investigated the strain anddisplacement response of the pile and temperature
distribution surrounding the pile considering the soil
domain as linear elastic and isotropic. Rajnee Saggu
and T. Chakroborty (IIT, Delhi) , 2014 studied
stresses and displacement of the energy piles in sandand the surrounding soil under cyclic thermomechanical loading” on abaqus simulia using
Abaqus Simulia. They applied 50 cycles of thermalloading of ∆T = 21°C on different load magnitudes
and concluded that Negative Shear stress isgenerated near the pile head in the soil and The
ultimate capacity (limit load) of the energy pile isunaffected by temperature changes.

Simulation and Parametric Studies
In this study, nonlinear transient thermomechanical axisymmetric FE analyses have been
carried out for floating energy piles in dense sands.
The pile and soil geometries are created as separate
parts using FE software Abaqus. Figure 1 showspile–soil geometry and the FE mesh for the pile–soil
domain. The vertical far-field soil boundary in themodel has been placed at a distance of fifteen timesof the pile diameter from the pile–soil interface. The
bottom soil boundary is placed at a distance of one
pile length from the bottom of the pile.
The far-field vertical soil boundary has been104

GEOMATE- Bangkok, Nov. 14-16, 2016

simulation results with the field pile load test dataand numerical simulation results given by Laloui etal. (2006).
Laloui et al. (2006) performed thermal pile load
tests in Lausanne, Switzerland for a duration of 28days in which heating period consisted of 12 days
and cooling period consisted of 16 days. The lengthand diameter of the pile were 26 m and 1 m,
respectively. The change in temperature was on theorder of ΔT = 21 °C induced in the pile. The pile
was placed in a layered soil deposit consisting of
alluvial soils, sandy gravelly moraine and mollase.
An axial load of 1,300 KN was applied at pile head.
Laloui et al. (2006) modelled the field pile load test
numerically using the Drucker–Prager thermoelastoplastic model for soil and linear elastic
constitutive model for pile. In the present simulation,
axisymmetric, thermo-mechanical FE analysis of a
26 m long pile with 1 m diameter is performed using
the coupled temperature-displacement procedure in
Abaqus. The elastic and thermal material properties
for the concrete pile and the soil layers have beentaken from Laloui et al. (2006) and reported inTable 3. The soil stress–strain response is simulatedusing Mohr–Coulomb model. The pile is assumed tobe thermo-elastic in Nature.

restrained in horizontal direction using roller support
and the bottom soil boundary has been restrained inboth vertical and horizontal directions using pinned
support. The thermal boundary conditions applied in
the model include heat flow through the far-fieldvertical and the bottom boundaries of the mesh. Heatflux is assumed to be zero along the axis ofsymmetry of pile and soil. The ambient soil
temperature is assumed to be 15 °C (Laloui et al.
2006). The FE domain has been discretized using
four node, axisymmetric, thermally coupled, bilinear
displacement and temperature elements (CAX4T)
for both soil and pile. At the pile–soil interface,shear strain localization may happen and as a result,shear band formation may take place in the firstcolumn of elements next to the pile shaft. Therefore,a refined mesh has been used near the pile–soilinterface in soil and coarser mesh in the far-fieldregions.
The interfaces between the pile and the soil
surfaces have been modelled as frictional contact inthe tangential direction with a coefficient of friction(μ) between soil and concrete pile = tanϕ where ϕ isthe internal friction angle of sand (saggu et al, 2012).
In the normal direction between the pile and the soil,
hard contact with zero penetration has been
considered.
Heat conduction between pile and soil has been
made possible by defining a linearly varying gap
conductance value. It is assumed that when the
distance between the pile and the soil is zero, thenthe conductance of contact between sand and pile isassumed to be perfect and this value reduces linearly
to zero for a gap size of 0.1m between the pile andthe soil. However, it is observed from the analyses
that no gap formation takes place between pile andsoil during thermal cycling. Thus, the gap
conductance at the pile–soil interface may beconsidered uniform and same as soil thermal
conductivity.

Properties

Soil
A1

Soil
A2

Soil
B

Soil
C

Soil
D

Pile

0.146

0.146

0.2

0.2

0.15

0.2

259

259

451

634

3865

27800

30°

27°

23°

27°

-

-

0

0

0

0

-

-

5000

3000

6000

2000
0

-

-

Thermalconductivity(k s ) (W/m °C)

1.8

1.8

1.8

1.8

1.1

2.1

Specific heat
(C s ) (J/°C)
Coefficient ofthermalexpansion(α s )(/°C)

1200

1230

1200

1091

785

800

10-5

10-5

10-4

10-4

10-6

10-5

MechanicalParameters
Poisson’s Ratio(υs)
Elastic
modulus (Es)
(MPa)
Mohr–
Coulombplasticity
parameters
Frictional angle(ϕ)
Dilation angle(ψ)
Apparentcohesion (c s )(Pa)
ThermalProperties

Validation of Finite Element Model underThermo-Mechanical Loading
For this study nonlinear transient thermomechanical axisymmetric FE analyses of Energy
Pile is performed on Software Abaqus Simulia. For
the checking the simulated programmes reliability
first a full scale in-situ pile was simulated and theresults were found to be in good agreement with thefield results. The Full scale test chosen for validation
was the test performed at EPFL by Lyesse Laloui as
the availability of data and extensive work done on
the Model.
The FE analysis results of geothermal energy
piles under combined mechanical and thermalloading have been validated by comparing theTable 1: Material Properties Used for First
Validation
105

GEOMATE- Bangkok, Nov. 14-16, 2016

The results of Simulation were found to be in good
agreement with experimental results.
A total of 10645 element are used for simulationof three different sizes with minimum being
0.005625 m2 in near regions to the pile andmaximum is 0.0675 m2 in far field regions while in
the finite element analysis by Laloui et al, coarse
634 elements were consider

Thus, the results of Simulation were found to be
in good agreement with experimental results.

Figure 3: Pile soil Geometry and the FE meshfor pile soil domain for present simulation.
A total of 3800 CAX4T elements were generated
with minimum size of 0.005625 m2 in near regionsof pile and maximum mesh size of 0.0675m2 in far
field regions. The material properties used for thesimulation and parametric studies are given as
follow:

Figure 1: EPFL energy pile-soil profile, Meshelements used for simulation (mesh elements nearpile element are kept finer and in far boundary
the mesh is coarser)

Properties

axial sterss in (MPa)depth of pile (in meters)-4.00E+00 -3.00E+00 -2.00E+00 -1.00E+00 0.00E+00
0Geothermal
Pile

0.2
70

0.2
27800

5

Mohr–Coulomb
parameters

10

Frictional angle (ϕ)

30°

-

15

Dilation angle (ψ)
Apparent cohesion (c s ) (Pa)
Thermal Properties

0
5000

-

Thermalconductivity(k s )(W/m °C)
Specific heat (C s ) (J/°C)
Coefficientofthermalexpansion (α s )(/°C)

1.8

2.1

1200

800

-5

10-5

20

validation
laloui exp.

Mechanical Parameters
Poisson’s Ratio (υs)
Elastic modulus (Es) (MPa)

Homogeneous
sand

25
laloui- numerical
30

plasticity

10

Table 2: Material Properties Used for Parametricstudies.

Figure 2: Comparison between Present
Simulation Results and Experimental andNumerical Simulation Performed by Laloui et al,
2006.

The thermo-mechanical analyses is done on a single
pile by for 6 load cases 50 KN, 250 KN, 500 KN,1000 KN, 1500 KN and 2000 KN for 3 thermalvariation in the pile ΔT = 15 °C, ΔT= 25 °C,
106

GEOMATE- Bangkok, Nov. 14-16, 2016

ΔT=35°C.Three aspects were mainly analysed in the
study, Axial Stress behaviour along the length ofpile, Radial Stress and Radial Strain Behaviour.
Results and Discussion:(a). At 50 KN Axial axial Load(d). At 1000 kN load

axial stress in Mpa
-0.8

-0.6

-0.4

axial stress in Mpa

-0.2

0

-1.5

0

at Temp cycle (15-30-15) celsius

at Temp cycle (15-30-15) celsius

at temp cycle (15-40-15)

at temp cycle (15-40-15)

at temp cycle 15-50-15

at temp cycle 15-50-15(e). At 1500 KN axial load

axial stress in Mpa

axial stress in Mpa

-0.5

0

-2

0

-1.5

-1

-0.5

0
0

5

5

10

10

15

15

20

20

at Temp cycle (15-30-15) celsius

at Temp cycle (15-30-15) celsius

at temp cycle (15-40-15)

at temp cycle (15-40-15)

at temp cycle 15-50-15

at temp cycle 15-50-15(c). At 500KN axial load(f). At 2000 KN axial load

axial stress in Mpa
-1

-0.5
0
5
10
15
20

(b). At 250 KN axial Load

-1

-1

0
5
10
15
20

axial stress in Mpa

-0.5

0

-3

0

-2

-1

0
0
5

5

10

10

15

15

20

at Temp cycle (15-30-15) celsius

at Temp cycle (15-30-15) celsius

at temp cycle (15-40-15)

at temp cycle (15-40-15)

at temp cycle 15-50-15

at temp cycle 15-50-15

Figure 4: Axial Stresses in Pile after Complete Heating Cooling Cycle at different loading(a). At 50 kN load (b). At 250 kN load (c). At 500 kN load (d). At 1000 kN load (e). At 1500 kN
load (f). At 2000 kN load
107

GEOMATE- Bangkok, Nov. 14-16, 2016

2. Axial Strain in Pile after Complete Heating and Cooling Cycle(d). At 1000 KN load

(A). At 50KN Loadaxial strain in 10 -6
-30

-20

-10

axial strain in 10^-6

0

-60

5
10
15
at temp var 15-50-15

-40

20

5
10
15
20
at temp var 15-50-15
at temp var (15-40-15)

at temp var (15-30-15)(e). At 1500 KN Load(b). At 250 KN Loadaxialstrain in-10
10-6
-30
-20

axial strain in 10^-6

0

-60

-40

5
10
15

10
15
at temp var 15-50-15 20
at temp var (15-40-15)
at temp var (15-30-15)(f). At 2000 KN Loadaxial strain in 10-6(c). At 350 KN Loadaxial strain in 10^-6
-10

-80

0

-60

-40

-20

5
10
15

0
0

depth of pile in meters

depth of pile in meters0

at temp var 15-50-15

05

at temp var 15-50-15 20
at temp var (15-40-15)
at temp var (15-30-15)

-20

-20
0

depth of pile in meters

depth of pile in meters

0

-30

0
0

at temp var (15-40-15)

-40

-20

depth of pile in meters

depth of pile in meters

0

5
10
15
at temp var 15-50-15

at temp var (15-40-15)

at temp var (15-40-15)

at temp var (15-30-15)

at temp var (15-30-15)

20

Figure 5: Axial Strain in Pile after Complete Heating Cooling Cycle at different loading(a). At 50 kN load (b). At 250 kN load (c). At 500 kN load (d). At 1000 kN load (e). At 1500 kN
load (f). At 2000 kN load
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GEOMATE- Bangkok, Nov. 14-16, 2016

3. Radial Strain in Pile after Complete Heating and Cooling Cycle(a). At 50 KN Load(d). At 1000 KN Loadradial strain in 10-6
-6

-4

-2

radial strain in 10^-6
0

2

-10

0

5
10
15
20
at temp var 15-50-15

1015
at temp var 15-50-15
20 var (15-40-15)
at tempat temp var (15-30-15)

at temp var (15-30-15)(b). At 250 KN Load(e). At 1500 KN Loadradial strain in 10-6
-4

-2

0

2

radial strain in 10^-64-5depth of pile in meters

depth of pile in meters

0
5
10
15
20
at temp var 15-50-15

0

10
15
20
at temp var 15-50-15
at temp var (15-40-15)(f). At 2000 KN Loadradial strain in10-6

radial strain in 10^-6
2

4

-10

6

0

depth of pile in meters

depth of pile in meters

15

at temp var (15-30-15)(c). At 500 KN Load

0

10

5

at temp var (15-30-15)

-2

5

0

at temp var (15-40-15)

-4

10

5

at temp var (15-40-15)

-6

5

0

depth of pile in meters

depth of pile in meters

0

-5

5
10
15
at temp var 15-50-15
at temp var (15-40-15)
at temp var (15-30-15)

0

10

20

0
10
20
at temp var 15-50-15
at temp var (15-40-15)
at temp var (15-30-15)

Figure 6: Radial Strain in Pile after Complete Heating Cooling Cycle at different loading(a). At 50 kN load (b). At 250 kN load (c). At 500 kN load (d). At 1000 kN load (e). At 1500 kN load(f). At 2000 kN load

109

GEOMATE- Bangkok, Nov. 14-16, 2016energy piles under different combinations ofthermo-mechanical loads different combinationsof thermo-mechanical loads, (February 2016).
http://doi.org/10.1016/j.compgeo.2015.11.010
Kramer, C., & Basu, P. (2014). Performance of amodel energy pile in sand. International
Conference on Physical Modelling in Geotechnics,(August2014),
1–6.
http://doi.org/10.1201/b16200-106
Laloui, L., Nuth, M., & Vulliet, L. (2006).
Experimental and numerical investigations of the
behaviour of a heat exchanger pile. International
Journal for Numerical and Analytical Methods inGeomechanics,
30(8),
763–781.
http://doi.org/10.1002/nag.499
Loria, A. F. R., Gunawan, A., & Ng, C. W. W.
(2015). Numerical modelling of energy piles insaturated sand subjected to thermo- mechanicalloads. Geomechanics for Energy and theEnvironment,
1(June),
1–15.
http://doi.org/10.1016/j.gete.2015.03.002
Loveridge, F. (2012). The Thermal Performance ofFoundation Piles used as Heat Exchangers in
Ground Energy Systems, 206.
Suryatriyastuti E., Mroueh, B. (2006). Thermomechanics behavior of energy pile subjected bymonotonic thermal loading Abstract : 20Ème
Congrès Français De Mécanique [Cfm2011],(2005), 4–9.
Saggu, R., & Chakraborty, T. (2014). Cyclic
Thermo-Mechanical Analysis of Energy Piles in
Sand. Geotechnical and Geological Engineering,33(2), 321–342. http://doi.org/10.1007/s10706014-9798-8
Marto, A., Teknologi, U., Mahir, A., & Universiti,
M. (2015). Comparisons on the response ofshallow geothermal energy pile embedded in softand firm soils Jurnal Teknologi C OMPARISONSON THE R ESPONSE OF S HALLOW,(February2016).
http://doi.org/10.11113/jt.v77.6433
Mimouni, T., & Laloui, L. (2014). Towards a securebasis for the design of geothermal piles. Acta
Geotechnica,9(3),355–366.
http://doi.org/10.1007/s11440-013-0245-4
Mimouni, T., Dupray, F., & Laloui, L. (2014).
Estimating the geothermal potential of heatexchanger anchors on a cut-and-cover tunnel.
Geothermics,
51,
380–387.

Summary and Conclusion
Geothermal Pile subjected to heating at different
loading, varying both thermal load and mechanicalload was analysed for a floating pile of diameter 1
meter and length 7.5 meter in dense sand.
The stress strain behaviour was analysed and it may
be concluded that• At low mechanical load the axial stress generated
on the pile varies much on applying thermalloading while when Mechanical Load is high, theaxial stress along the depth varies lesser. So it may
be concluded that when mechanical load value is
low the thermal variation influences the axial
stress produced and at higher mechanical load thestress behaviour is nearly constant at different
temperature variation, so the axial stress along thepile is majorly due to applied mechanical load.
• The radial strain in the pile was also influenced
only at low mechanical load, at higher loads thevalue of radial stress and strain is nearly same.
• Larger axial strain were noted at the tip of pile andsand surrounding the energy pile near head of pile.

References:
Abaqus/Standard User’s Manual, Version 6.11
(2011) Dassault Syste`mes Simulia Corporation,
Providence, Rhode Island, USA.
Amis, A., Bourne-Webb, P., Amatya, B., Soga, K. &Davidson, C. (2008), "The effects of heating andcooling energy piles under working load at
Lambeth College". Proc. 33rd Ann. and 11th Int.
DFI Conf., New York, article no. 1620.
AMATYA, & Laloui, L. (2012). Thermomechanical behaviour of energy piles, (6), 503–
519.
Arson, C., Berns, E., Akrouch, G., Sanchez, M., &Briaud, J. (2013). Heat Propagation around
Geothermal Piles and Implications on Energy
Balance, 628–635.
Brandl, H. (1998). “Energy piles and diaphragmwalls for heat transfer from and into the ground."Proc. 3rd Int. Symp. on Deep Foundations on
Bored and Auger Piles, BAP III, Ghent, 37–60.
Brandl, H. (2006). "Energy foundations and otherthermo-active ground structures". Ge´otechnique
56,
No.
2,
81–122,
http://
dx.doi.org/10.1680/geot.2006.56.2.81.
Bourne-Webb, P., Amatya, B. L., Soga, K., Amis, T.,Davidson, C. & Payne, P. (2009). "Energy pile
test at Lambeth College, London: geotechnicaland thermodynamic aspects of pile response toheat cycle". Ge´otechnique 59, No. 3, 237–248,
http:// dx.doi.org/10.1680/geot.2009.59.3.237.
Donna, A. Di, Torino, P., Loria, A. F. R., Donna, A.
Di, Loria, A. F. R., & Laloui, L. (2015).
Numerical study of the response of a group of

http://doi.org/10.1016/j.geothermics.2014.02.007
Ng, C., Shi, C., Gunawan, A., Laloui, L., & Liu, H.
L. (2015). Centrifuge modelling of heating effects
on energy pile performance in saturated sand.
Canadian Geotechnical Journal, 1057(December2014), 1045–1057. http://doi.org/10.1139/cgj2014-0301

110SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

MODEL TEST ON THE REINFORCING EFECT OF THE FACE BOLTS
IN THE UNCONSOLIDATED SANDY LAYER TUNNEL
Yoshifumi Taguchi1
1

Department of Comprehensive Engineering, Urban Environment Course, Kindai University TechnicalCollege, Japan

ABSTRACTIn the tunnel excavating unconsolidated sandy layer in NATM, it is important to stabilize the tunnel cutting
face and to prevent the surface settlement. Therefore, various auxiliary methods of tunnel construction are usedby this purpose, but it becomes the effective method to drive the long face bolts recently. As the effect of theface bolts for the cutting face stability and the surface settlement control, it is thought that reinforcing effects aredifferent according to length and the interval. In this study, we changed the length and the interval of the facebolts by the two-dimensional model test used Toyoura-sand and examined the influence level on the effect of theface stability and the surface settlement. As a result, face stabilizing effect is higher better installation interval of
the face bolt is small. When the length of the face bolt is longer than 0.3D（D: cutting face height）, thereinforcing effect of the cutting face becomes higher. It is small in length 0.2D. When the length of the facebolt is longer than 0.5D, about the surface settlement, a reinforcing effect becomes higher. In addition, theinfluence range of the surface settlement is 1.3-2.0D before from the tunnel cutting face.
Keywords: Tunnel, Face bolt, Model test, Unconsolidated sandy soil, Reinforcement effect

INTRODUCTION

and 15cm in width.
The actual tunnel is a three-dimensional crosssection of a circular, but the reason for the twodimensional, a scale of the soil tank becomes large
in three-dimensional tunnel model, the test becomesdifficult, but the two dimensional section is because
easy to do experiments or analysis.
In this test, a cutting face bolt modeled with athin Kent paper thicknesses 0.12mm of a flat plate
shape, was placed at a predetermined interval in
cutting face, as shown in Fig. 3.

In the tunnel excavating unconsolidated sandy
layer in NATM as shown in Fig.1, it is important tostabilize the tunnel cutting face and to prevent thesurface settlement. Therefore, the various auxiliary
methods of tunnel construction are used by thispurpose, and it becomes the effective method todrive the long face bolts recently.
As the effect of the face bolts for the cutting facestability and the surface settlement control, it isthought that reinforcing effects are different
according to length and the interval. In this study,
we changed the length and the interval of the facebolts by the two-dimensional model test usedToyoura-sand and examined the influence level on
the effect of the face stability and the surface
settlement[1]-[6].

Building
House

Ground Surface

OUTLINE OF MODEL TEST
Sandy

Model Test Equipment

soil

Shotcrete andsteel support

A model test equipment shown in Fig. 2. The
size of the soil tank is 65cm in height, 80cm in width,and 15cm in depth. As shown in Fig. 3, Tunnel partof the model is the two-dimensional cross section of
the longitudinal direction taken along the center of
the actual tunnel cross section, a rectangular crosssection of the cutting face model is 15cm in height

TunnelCutting faceLong face bolt

Fig.1 MATM tunnel in unconsolidated ground(face bolt combination)
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GEOMATE-Bangkok, Nov. 14-16, 2016

Front

overburden thickness 30cm (2.0H). The density ofthe sand is 1.54g/cm3.
The pull out of the tunnel, one by 0.2mm up to 0～2mm, one by 0.5mm up to 2～10mm, one by
1mm up to 10 ～ 15mm was carried out. The
measurement of the surface settlement used a
displacement gauge. It was placed in five locationsin the 10cm interval.

Side
view

view

Displacement
gauge

Sand

Sand
Cutting face

Table 1 Case of model test
Load cell

Model testcase

Fig.2 Model test equipment

Packing

Length（ℓ）

1Face bolt model (Kent-paper)

Installationinterval

Non reinforced

2

15.0cm (1.0H)

1,3,5-stage

3

7.5cm (0.5H)

1,3,5-stage

4

3.0cm (0.2H)

5-stage

Face
Pull-outby jack

Supportplate

EXPERIMENTAL RESULT

Load cell

Case of Non Face Bolt
Fig.3 Cutting face part enlarged view (unit mm)

Figure 4 shows the change of the load acting on
the tunnel cutting face due to the pull out of the
cutting face model when there is no face bolt.
From this figure, the load acting on the cutting facedecreases rapidly in the pull-out of about 1.0mm,
then, it converges to 4～5N. This may be because
the earth pressure of the cutting face is made fromthe earth pressure at rest to active earth pressurestate.

Experimental Method

Load on cutting face (N)The model test reproduced the excavation bypulling out the tunnel cutting face model by a screw
jack. Upon reaching a predetermined pull-out
amount was measured the load acting on the tunnel
cutting face and the surface settlement. The load
acting on the tunnel cutting face sets the load-cell on
the back side of the tunnel cutting face, wasmeasured change of the load in accordance with thepull-out.
Table 1 shows the case of the model test. The
length of the face bolt model (Kent paper) waschanged 15cm (1.0H, H: cutting face height ofmodel tunnel), 7.5cm (0.5H) and 3cm (0.2H).
Further, the installation interval was changed 7.5cm(1- stage), 5cm (3 -stage), 3cm (5- stage), andinvestigated the influence.
The model of the face bolt used a flat plate shapeof Kent paper instead of rod-type bolt is because
easy simply compare the effect of the length orinterval of the bolts, also because the model
preparation near the tunnel cutting face is easy.
Model ground was used Toyoura-sand of the airdried state. The model sand was free-fall from theoutlet of a fixed size at the position of the height
60cm, and the model ground was introduced to◇ Non-reinforced

Pull-out amount (mm)Fig.4 Load on cutting face due to pull-out(non-reinforced)
Similarly, it shows the change in surface
settlement due to tunnel cutting face pull out in Fig.5.
The measurement point just above the face, andfront 10cm, 20cm, 30cm and 40cm from the cutting

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GEOMATE-Bangkok, Nov. 14-16, 2016

Surface settlement (mm)Pull-out amount (mm)Figure 7 shows the change in surface settlement
due to cutting face pull-out. The settlement
measurement points are the same as described above
the non-face bolt case. The surface settlement
amount of point of the face above and face the front
10cm is slightly larger. The maximum surface
settlement at the time of pull-out 15mm of tunnel
face is about 2 mm, surface settlement than non-facebolt is considerably suppressed.
Pull-out amount (mm)Surface settlement (mm)face just above. It is the point of the rear 10cm fromthe cutting face just above. The surface settlement of
the point of the cutting face immediately above andthe front 10cm is large. With the pull-out amount of
the cutting face increases, the surface settlement isincreasing linearly.
The reason for this surface settlement has
increased because its position is within the range ofthe sliding surface of the ground. The surface
settlement amount at the position of the cutting facebehind 10cm and face forward 20cm~40cm is small
and the surface settlement is not significantly
increased even if the pull-out amount of the cutting
face is increased. The reason is because the surface
settlement is not significantly increased in order tothat position the outside than the range of the sliding
surface.

Face backward 10cm
Face just above
Face forward 10cm
Face forword 20cm
Face forword 30cm
Face forword 40cm

Fig.7 Surface settlement due to pull-out(Bolt length 15cm, 5- stage)
Face backward 10cm
Face just above

Influence of Number of Bolts on the Load Acting
Cutting Face

Face forward 10cm
Face forword 20cm
Face forword 30cm
Face forword 40cm

In the case of length 15cm (1.0H)

Fig.5 Surface settlement due to the pull-out(non-reinforced)

Figure 8 shows the comparison of the differencein load acting on the cutting face due to thedifference of the face bolt installation interval in thecase of face bolt’s length 15cm (1.0H). Installationinterval of the face bolt is three types of 1-stage, 3stage and 5-stage. In all cases, the cutting face loadin the pull-out of about 1.0mm is reduced rapidly,
has converged constant value as the pull-out amountincreases. In the case of the face bolt 5-stage, cutting
face load has become the smallest, the facestabilizing effect is higher. In the case of 3-stages,
the cutting face load has become slightly larger thanin the case of 5-stages, it is substantially equal.

Case of Using Face Bolts

Load on cutting face (N)Load on cutting face (N)Figure 6 shows the change in the load acting on
the cutting face due to the tunnel face pull-out in thecase of the face bolt installed 5-stage at the
installation interval 3cm and length 15cm. From thisfigure, the load acting on the cutting face decreases
rapidly in the pull out of about 1.0mm, then, it
converges to about 2N. This value is smaller thanthe value of non-reinforced case, have appeared
reinforcing effects of the cutting face bolts.

◇ L=15cm, 5-stage□

Non-reinforced△

L=15cm, 1-stage○

L=15cm, 3-stage◇

L=15cm, 5-stage

Pull-out amount (mm)Pull-out amount (mm)Fig.8 Load on cutting face due to pull-out(Bolt length 15cm (1.0H))

Fig.6 Load on face due to pull-out(Bolt length 15cm, 5-stage)
113

GEOMATE-Bangkok, Nov. 14-16, 2016

Even if the face bolt 1-stage, the face load issmaller than the case of non-face bolt in the cutting
face pull-out early stage, the face stabilizing effect isobserved. However, as the pull-out amount of theface is increased, it has become almost equivalencein the case of non-face bolt, the reinforcing effect issmall.

become the smallest, the face stabilizing effect ishigher.
On the other hand, in the case of the length of
3cm, the face load is smaller than in the case of nonreinforced in the face pull-out early stage. However,as the pull-out amount of the face is increased, it has
become almost equivalence in the case of nonreinforcing, the reinforcing effect is small. In thecase of the length 7.5cm it is almost equal to that of
the length 15cm.

In the case of length 7.5cm (0.5H)

Load on cutting face (N)Figure 9 shows a comparison of differences ofload acting on the cutting face due to the differencein the installation interval of the face bolt when thelength 7.5cm face bolt (0.5H). Installation interval of
the face bolt is three types of 1-stage, 3-stage and 5stage. In all cases, the load acting on the face rapidly
decreases with the pull-out of about 1.0 mm, isconverged to a constant value with the pull-out
amount is increased.
In the case of face bolt 5-stages, the load acting
on the face has become the smallest, high facestabilizing effect. In the case of the 3-stages, thisface load is larger slightly than in the case of the 5stages. In the case of face bolt 1-stage is almost thesame tendency as in the case of face bolt 15cm (1stage) of the above-mentioned. In the case of facebolt 5-stage, the difference due to the length 15cm
(1.0H) and 7.5cm (0.5H) in length it is not seen so
much.

□

Non-reinforced△

L=15cm, 5-stage○

L=7.5cm, 5-stage◇

L=3.0cm, 5-stage

Pull-out amount (mm)Fig.10 Load on cutting face due to pull-out(Bolt installation interval 5-stages)
Influence of the Installation Number of Face Bolt
to Surface Settlement

Load on cutting face (N)In the case of bolt length 15cm
Figure11 to 13 shows the surface settlement of
the tunnel longitudinal direction in the case ofchanging the installation interval of the face bolt
when the case of face bolt length is 15cm (1.0H).
The tunnel cutting face position as 0cm, at theposition of the face front 10cm, 20cm, 30cm, 40cm
and face backward 10cm, shows the surface
settlement value of each case.
Figure 11 shows the surface settlement whenpulled out 5mm (0.03H) the cutting face. Although
in the case of non-reinforcement are about 2mm
settlement at the point of the face front 10cm fromthe face just above the part, hardly settlement in theface bolt 5-stage and 3-stage. In the case of the facebolt 1-stage, slight settlement occurs.

□ Non-reinforced△ L=7.5cm, 1-stage○ L=7.5cm, 3-stage◇ L=7.5cm, 5-stage

Pull-out amount (mm)Fig.9 Load on cutting face due to pull-out(Bolt length 7.5cm (0.5H))
Influence of Length on the Load Acting Cutting
Face

Distance from tunnel cutting face (cm)Surface settlement (mm)Figure 10 shows the difference in load acting on
the cutting face due to the difference in length of theface bolts when the installation interval of the facebolt is 5-stages. The length of the face bolt is three
types of 15cm, 7.5cm and 3cm. In all cases, the load
acting on the face rapidly decreases with the pull-out
of about 1.0 mm, is converged to a constant value
with the pull-out amount is increased. In the case of
the length of 15cm, the load acting on the face

□ Non-reinforced△ L=15cm, 1-stage○ L=15cm, 3-stage◇ L=15cm, 5-stage

Fig.11 Surface settlement curve(Longitudinal direction, on 5mm pull-out)
114

GEOMATE-Bangkok, Nov. 14-16, 2016

Similarly, Fig.12 shows the surface settlement
when the cutting face was pulled out 10mm (0.07H).
In the case of non-reinforced, at the point of the facefront 10cm from the face above, it is about 4mm
settlement. On the other hand, in 5-stage and 3-stage
hardly subsidence, about 1mm settlement in the face
just above even 1-stage, the surface settlement
control effect is clearly evident.

Figure 15 shows the surface settlement whenpulled out 15mm (0.1H) . In the case of nonreinforcement are about 6～7mm settlement at the
point of the face front 10cm. On the other hand, inthe 5-stage and 3-stage is a settlement of about 2mm,
a settlement of about 3mm in the 1- stage.

Surface settlement (mm)Distance from tunnel cutting face (cm)Surface settlement (mm)Distance from tunnel cutting face (cm)□ Non-reinforced△ L=15cm, 1-stage○ L=15cm, 3-stage◇ L=15cm, 5-stage○ L=7.5cm, 3-stage◇ L=7.5cm, 5-stage

Surface settlement (mm)Distance from tunnel cutting face (cm)□ Non-reinforced△ L=7.5cm, 1-stage○ L=7.5cm, 3-stage◇ L=7.5cm, 5-stage

Fig.15 Surface settlement curve(Longitudinal direction, on 15mm pull-out)
Influence of the Length on the Surface Settlement

Distance from tunnel cutting face (cm)Surface settlement (mm)△ L=7.5cm, 1-stage

Fig.14 Surface settlement curve(Longitudinal direction, on 5mm pull-out)

Fig.12 Surface settlement curve(Longitudinal direction, on 10mm pull-out)
Figure 13 shows the surface settlement whenpulled out 15mm (0.1H) the cutting face. In the caseof non-reinforcement is about 6～7mm settlement at
the point of the face front 10cm. Although the 5stage and 3-stage have been settlement of about
2mm, increased sharply settlement just above thecutting face in the case of 1-stage, there is no
reinforcing effect. In addition, from these figures,the range of influence of surface settlement is aboutface forward 20 ~ 30cm (1.3 ~ 2.0H).

□ Non-reinforced

Figure 16 to 17 shows the surface settlement of
the tunnel longitudinal direction changing the facebolt length 15cm (1.0H), 7.5cm (0.5H) and 3cm(0.2H) in installation interval 5-stages.
Figure 16 shows the surface settlement whenpulled out 5mm (0.03H) the cutting face. Although
in the case of non-reinforcement are about 2mm
settlement at the point of the face front 10cm fromthe face just above the part, hardly settlement in thecase of installing face bolts.

□ Non-reinforced△ L=15cm, 1-stage○ L=15cm, 3-stage◇ L=15cm, 5-stage

Fig.13 Surface settlement curve(Longitudinal direction, on 15mm pull-out)
In the case of bolt length 7.5cm

Distance from tunnel cutting face (cm)Surface settlement (mm)

Similarly, Fig.14 to 15 shows the surface
settlement of the tunnel longitudinal directionchanging the installation interval of the face bolt in
the case of face bolt length 7.5cm (0.5H). Figure 14
shows the surface settlement when pulled out 5mm(0.03H) the cutting face. Although in the case of
non-reinforcement is about 2mm settlement at the
point of the face front 10cm from the face just above
the part, in the case of face bolt 5-stage and 3-stage
is less than the half. In the case of the face bolt 1stage, the settlement is about 1mm.

□ Non-reinforced△ L=15cm, 5-stage○ L=7.5cm, 5-stage◇ L=3.0cm, 5-stage

Fig.16 Surface settlement curve(Longitudinal direction, on 5mm pull-out)

115

GEOMATE-Bangkok, Nov. 14-16, 2016

Similarly, Fig.17 shows the surface settlement
when pulled out 15mm (0.1H) the cutting face. In
the case of non-reinforcement are about 7～8mm
settlement at the point of the face front 10cm.
Although the settlement amounts of the same degreein the face bolt length 15cm and length 7.5cm areabout 2mm settlement, the surface settlement control
effect of is high. On the other hand, in the face bolt
length 3cm (0.2H), the settlement amount has been
increased to about 4mm just above tunnel face, thesurface settlement control effect is small.

settlement control effect is large, settlement isincreased in 1-stage.
(4)When the length of the face bolt is longer than
0.5H, the surface settlement control effect of
becomes higher. In the installation interval is 5stage, although the length of 15cm (1.0H) and7.5cm (0.5H) are almost the same settlement
control effect, is large settlement in bolt’s length
3cm (0.2H).
REFERENCES[1]The Japan Geotechnical Society, “Toconstruction from the design of the mountaintunnel construction method”, 2007, pp.140-141.
(in Japanese)[2] Japan Society of Civil Engineers,”Practice ofmodel experiments and numerical analysis in themountain tunnel”, 2006,pp.22-59. (in Japanese)[3]Fukusima S, Mochizuki Y,Kagawa K, Yokoyama
A, “Model test on pre-reinforcement of shallow
tunnel in sandy ground”, Japan Society of Civil
Engineers, No.406, Ⅲ -11 ， JUN.1989,pp.7986.(in Japanese)[4] Taguchi Y, Kagawa K, Sagara M, Yoshikawa K,“ Reinforcing effects of thin flexible prelining”, Japan Society of Civil Engineers, No.
645,Ⅲ-50,Mar.2000, pp.125-135. (in Japanese)[5]Mitarasi Y, Matsuo T,Tezuka T,Okamoto
T,Nisimura S,Matsui T,”Evaluation of effect oflong face reinforcement method in tunneling”,
Japan Society of Civil Engineers, No.743,Ⅲ-64，
SEP.2003,pp.213-222.(in Japanese)[6]Domon T, Seo K, Nishimura K, ”Simplified
design method for face bolting based on thereinforcing effect”, Journal of Japan Society of
Civil Engineers, Ser. F1,Tunnel Engineering, vol.
66, No. 1,NOV.2010,pp.29-39.(in Japanese)[7] Tanaka T, Taguchi Y, ” Model Test on
reinforcing effect of the long face bolt in
unconsolidated sand”, Chubu branch Japan
Society of Civil Engineering,Mar.2010, pp 297298. (in Japanese)[8] Taguchi Y, Tanaka T,” Model test on the effectof reinforcing the face bolt in unconsolidated
sand tunnel”, Kindai University TechnicalCollege Research Reports, vol.4, JAN.2011,
pp.79-84. (in Japanese)[9] Taguchi Y, Tabira T,”Model test on tunnel facestabilizing effect of the face bolt”, Kindai
University Technical College Research Reports,
vol.6, MAR.2013, pp.71-76. (in Japanese)

Surface settlement (mm)Distance from tunnel cutting face (cm)□ Non-reinforced△ L=15cm, 5-stage○ L=7.5cm, 5-stage◇ L=3.0cm, 5-stage

Fig.17 Surface settlement curve(Longitudinal direction, on 15mm pull-out)
CONCLUSION
In this paper, by the two-dimensional
longitudinal direction of the model test using
Toyoura-sand, by changing the installation interval
and length of the tunnel cutting face bolt, examinedsurface settlement controlling effect and tunnel facestabilizing effect. As a result, it was found that thefollowing.
(1)Small installation interval of the face bolt andmore the number of installations is often thesmaller the load acting on the face, the facestabilizing effect becomes higher. In the bolt
length 15cm (1.0H, H: face height), the facestabilizing effect is higher in the installationinterval 5-stage and 3-stage, but it is small in the
1-stage.
(2)When the length of the face bolt is longer than
0.5H (H：face height), the load acting on theface is small, the reinforcing effect is increased.
In the case of installation interval 5-stages, inbolt’s length 15cm (1.0H) and 7.5cm (0.5H)
the face stabilizing effect is high, the bolt’s
length of 3cm (0.2H) is small.
(3)When the number of installations of the face bolt
is larger, the surface settlement control effect ishigher. In the case of bolt length 15cm (1.0H),
although in the 5-stage and 3-stage surface

116Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

TUNNEL LINING RESPONSES DUE TO ADJACENT LOADED PILE –NUMERICAL INVESTIGATION
N. Heama1, P. Jongpradist2, P. Lueprasert1 and S. Suwansawatt1
Faculty of Engineering, King Mongkut’s Institute of Technology Ladkrabang, Bangkok, Thailand
2
Faculty of Engineering, King Mongkut’s University of Technology Thonburi, Bangkok, Thailand

1

ABSTRACT
Underground structures are popularly utilized in urban development, especially tunnels for both transportation
and utilities. The interaction problem between existing tunnel and piles from new constructed structures thus
becomes unavoidable in dense area. The tunnel would deform or even lose its stability if the additional forces (dueto adjacent piles under loading) in lining are drastically high. This depends on many factors, such as the clearance
and pile tip level with respect to tunnel position. For preliminary assessment during the first stage of design, it isnecessary to estimate this impact. The concept of tunnel protection zone is commonly adopted. However, toestablish the tunnel protection zone for adjacent loaded pile, the understanding on this interaction problem isessential. This study analyzes the effect of adjacent pile under loading on the existing tunnel by 3D finite element
method. The case study is the tunnel of MRTA project subjected to an adjacent 1 m bored pile under loading withvarious lengths and clearances. The additional forces (bending moment and axial force) are investigated. The results
show that the additional forces increase when the tunnel affected by adjacent loaded pile. the maximum value
occurs at invert and right spring line.
Keywords: 3D FEM, Tunnel lining deformation, Structure forces, Adjacent pile

The numerical and finite element analyses have also
been utilized in design of tunnel lining [6-8].
Although the current design practice for tunnel
lining considers all possible load scenarios duringservice of the tunnel, however, the future activities
have not been included. It is then necessary toevaluate the possible effect of adjacent pile underloading on tunnel lining in terms of additional
structure forces. The appropriate mean can then be
prepared to tackle this problem.
Therefore, this study investigates the effects ofadjacent pile under loading on existing tunnel interms of the structural forces (bending moment andaxial force) by 3D FEM program.

INTRODUCTION
The underground structures increase in urban
development,
especiallytunnelsforbothtransportation and utilities. The Mass Rapid Transit
Authority of Thailand (MRTA) has used the tunnel tosolve the traffic congestion problems in Bangkok,
Thailand. Generally, the MRT tunnels are aligned
below major road and pass indirectly under buildings.
In recent years, numbers of building constructions are
increasing. Some new construction projects using a
deep pile foundation are close to existing tunnels. The
new constructed adjacent piles under loading as
shown in Fig.1 may induce the adverse effect ontunnel lining.
In recent years, 3D finite element method (FEM)
was used to analyze the effect on existing tunnel due
to loaded bore pile in London [1], the effect of surface
construction of multi-storey commercial building on
two existing tunnels of Toronto subway in Canada [2]and the effect of a deep open face excavation on
existing tunnel in Prague-Czech Republic [3]. In
Bangkok, Thailand, the new influence zone
considering the pile tip position with respect to tunnel
position was presented by [4].
However, those researches focused only on thetunnel deformation and did not consider the
additional forces in lining. Only a few works
considered the effect on the structural forces in lining.
The 2D and 3D FEM are used to study the structuralforces during tunnel excavation or even additional
forces in tunnel lining affected by groundstratification, surface buildings and tunnel depth [5].

Fig. 1117

New construction adjacent to existing tunnel
in urban environment.

GEOMATE- Bangkok, Nov. 14-16, 2016

METHODOLOGY
Fig. 2

Geometric parameters, the tunnel and boredpile installed.
Numerical model

Considered Problem and Variables
The evaluation of the effects of adjacent single
pile on existing tunnel is accomplished with the 3Dfinite element program- PLAXIS3D. Undrained
conditions were used in this study. The analysis did
not consider the influence of pile construction. The
length and applied working load of the single pile arevaried in the analysis. The working loads are derived
using the α – method [9]. In this paper, the geometricparameters in the study case are depicted in Fig. 2a.
The tunnel diameter ( DT) of 6. 3 m, the lining
thickness ( T) of 0. 3 m and the tunnel depth ( LT) of
20. 0 m below the ground surface, the bored pile
dimeter (DP) of 1.0 m and the interface of 0.9 which isinteraction between soil and pile elements wereconsidered. In addition, two case studies are taken
into account. The first case (CASE 1) considered thetunnel located in soft clay. The second case ( CASE
2) , the tunnel is located in stiff clay. In the analysis,
the clearance ( C) between the bored pile to edge oftunnel and the bored pile length ( LP) are varied asshown in Fig 2b.

The 3D-finite element mesh and numericalmodeling in this paper are shown in Fig.3. The
dimension of model is 80 m ( ≈ 12.5DT) in thetransverse direction, 60 (≈ 9.5DT) in the longitudinal
direction and 60 m (≈ 9.5DT) in the vertical direction.
The monitoring plane was assigned at the center of
longitudinal direction. To directly obtain the
structural forces, shell elements are used to model
tunnel lining with grouting layer, which is obtained
from [10, 11]. The detail of this technique ispresented in the companion paper [12].

Fig. 3Model and meshes for tunnel construction insoft clay.

Boundary condition and initial conditions
Boundary conditions
Displacement boundary conditions are used insimulation of this study. The left and right side andbottom boundaries are sufficiently extended from theeffect of tunnel excavation and pile settlement in thethree dimensional analysis. The sides of the meshincluding rear side and front side are free to movevertically but restrained against lateral movements.
Left and right sides of mesh are controlled against
perpendicular movement. The bottom of the mesh is
fixed to both horizontal and vertical movement. The
surface is allowed for free horizontal and verticalmovement.
Initial conditions
The initial stress state, horizontal effective stressand vertical effective stress are controlled by thecoefficient of earth pressure at rest, Ko and unit weightof soil layers. A typical pore water pressure profile
from case study in MRTA Blue Line Project inBangkok, Thailand in term of piezometric drawdown[7] with a water table at depth of 2.0 m below groundsurface.

118

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 4Model parameters
The linear elastic model is assumed for tunnel
lining, pile and grouting layer. The soft and stiff clay
is modeled by hardening soil model (HS) [13]. The
properties of soil are determined form MRT projects[14]. The material parameters for the simulations ofgeotechnical work in Bangkok subsoil are calibratedby [15].

Based on previous studies on impact of adjacentloaded pile on existing tunnel [5,7], the tunnel
deformation is generally considered in term of thechanged tunnel diameter on the tunnel axis withvertical (tunnel crown and Invert) and horizontal(tunnel spring line) directions. Therefore, themaximum bending moment and axial force were
measured from the corresponding positions as shownin Fig 4. The considering method of the change ofbending moment and axial force are shown in Eq. (1)and (2). When 𝟏 and 𝟐 are structural forces intunnel lining before and after pile under loading,respectively.

Table 1 Soil model parameters [16]Soil layer

Soft clay

Stiff clay

Material model

HS

HS

ref
(kPa)
Eoed

5,000

60,000

(kPa)

5,000

60,000Eurref (kPa)

15,000180,000 sat (kN/m2)

16

18

ref
50

E

'

(-)

0.33

0.33

'

( o)
(kPa)

22

22

5

18

m (-)

1

1

pref (kPa)

100

95

c

Positions to observe the structural forces intunnel lining.

M = change of bending moment
𝑴 = 𝑴𝟐 − 𝑴𝟏(1)

𝑵 = 𝑵𝟐 − 𝑵𝟏(2)

N = change of axial force

ANALYSIS RESULT
Structural forces in tunnel lining
The structural forces (bending moment and axial
force) of tunnel ling induced by adjacent pile underloading are observed and presented in this section.

Table 2 Material properties of the bore pile, tunnel
lining and grouting layer.
Youngmodulus[E](kN/m2)

Poisson’s
ratio
[ ]

Unit
weight[ ](kN/m3)

Bored
pile

3.1 x 107

0.20

24

Tunnellining3.1 x 107

0.20

24

Grouting
layer

1 x 106

0.30

21

Bending Moment
Figure 5 depicts the bending moment at crown,
invert and spring line in non-dimensional form when
the tunnel is located in soft clay and the clearances
between the pile and the tunnel are 0.5 and 4.5 m. The
values shown in the figure are normalized by the
initial bending moment (Mi) values (after completionof tunnel construction) at their positions. When thepile tip is above the tunnel axis, the bending moment
gradually increases with pile length and the maximum
value occurs when the pile tip is at the depth of 0.7075LT (0.75-0.5DT above the spring line). The
maximum values happen at the invert with the valueof 35% of the initial moment. After that, the moment
gradually decreases with increasing pile length. The
minimum value occurs at a depth in the range of 1.01.25LT. When the pile tip extends to the depth of
1.25LT to 1.75LT, the bending moment becomesincreasing again and gradually decreases until
constant or remains unchanged with extending pile tip
position underneath this level. Note that the values ofchange of bending moment in case of clearance of 4.5
m are much smaller than those in case of clearance of
0.5 m.

Measurement of Structural forces in lining

119GEOMATE- Bangkok, Nov. 14-16, 2016

moment gradually increases and the maximum value
occurs at a depth of 0.8LT, where the pile tip
penetrates into the stiff clay layer. After that the
moment decreases until the pile tip is located at depth
in the range of 1.0-1.25 LT before increasing with pilelength again. However, the ratios of increase of
moment (M/Mi) in this case is much smaller thanthose of the previous case (tunnel in soft clay). Notethat the maximum value appears at the crown for thiscase.
CASE 1, the maximum increments of bending
moment at crown and invert increase 8- 10 and 5- 35
percent, respectively. While at the left and right sidesof spring line, 3- 5 and 10- 25 percent of maximum
additional bending moment are observed.
CASE 2, the maximum increments of bending
moment at crown and invert increase 10- 20 and 1055 percent, respectively. While at the left and rightsides of spring line, 5- 10 and 8- 50 percent of
maximum additional bending moment are observed.

Fig. 5The change of bending moment in tunnel
lining due to adjacent loaded pile in softclay.

Fig. 7The change of bending moment in tunnel
lining at invert due to adjacent loaded pile insoft clay.

To investigate the effect of clearance on induced
bending moment, the change of bending moment intunnel lining due to adjacent pile with various
clearance are shown in Fig.7. Only the case of tunnel
located in soft clay and at the invert is shown. Thechange of bending moment decreases when clearance
increases. From the figure, it is seen that only smallchange in tunnel lining moment occurs if theclearance is larger than 3.5 m. This agrees well withthe previous study [7] on tunnel lining deformation.
The clearance of 3.5 m seems to be the safe distancefor this condition.
Fig. 6

Axial Force

The change of bending moment in tunnel
lining due to adjacent loaded pile in stiffclay.

Figure 8 depicts the axial force at crown, invert
and spring line in non-dimensional form when thetunnel is located in soft clay and the clearances
between the pile and the tunnel are 0.5 and 4.5 m. The
values shown in the figure are normalized by the

Figure 6 depicts the analysis results when thetunnel is located in stiff clay. With increasing depthof the pile tip above the tunnel axis, the bending120

GEOMATE- Bangkok, Nov. 14-16, 2016

initial axial force (Ni) values (before having a loadedpile) at their positions. When the pile tip is above thetunnel axis, the axial force dramatically increases
with pile length and the maximum value occurs when
the pile tip is at the depth of 0.8LT (0.5DT above thetunnel spring line). Unlike the bending moment, themaximum value occurs at the right side of spring line.
After that, the axial force gradually decreases withincreasing pile length. The minimum value occurs at
a depth in the range of 1.0-1.25LT. When the pile tip
extends to the depth below the tunnel axis, the axial
force becomes constant or remains unchanged in all
clearance. Note that the values of change of axial
force in case of clearance of 4.5 m are much smaller
than those in case of clearance of 0.5 m. In CASE 2(tunnel located in stiff clay), the ratios of increase of
axial force (N/Ni) is much smaller than those of the
CASE 1.
CASE 1, the maximum increments of axial
force at crown and invert increase 1- 2 and 0. 5- 1
percent, respectively. While at the left and right sidesof spring line, 0. 5- 1 and 1- 4 percent of maximum
additional bending moment are observed.
CASE 2, the maximum increments of axial
force at crown and invert increase 1- 7 and 0. 5- 1
percent, respectively. While at the left and right sidesof spring line, 5- 10 and 0. 5- 10 percent of maximum
additional bending moment are observed.

By comparing with the result form previous study[17] on tunnel deformation as shown in Fig.9. The
position of pile tip to induce maximum deformationagrees well that in this study.

Soft Clay

Stiff ClayØH2ØH1ØV2 ØV1

Fig. 9The change of tunnel diameter due toadjacent loaded pile in stiff clay. (modified
from [17])CONCLUSION
This paper analyzes the effect of existing tunnel
due to adjacent single pile under loading by 3D finite
element method. The single pile is positioned withvarious lengths and clearances. The changes ofstructural forces in tunnel lining are observed. The
main results of the analysis are as follows:
1. The structural forces in tunnel lining that is
located in stiff clay are larger than those oftunnel located in soft clay. This is due to thatthe working load on pile in stiff clay ( CASE
2) is larger than that in soft clay (CASE 1).
2. The structural forces decrease whenclearances
increaseandbecomeinsignificantly increasing when the clearance
is larger than 3.5 m. The maximum structuralforces in tunnel lining occur when the pile tip
is located at the depth of about 0. 8LT in all
clearance and all cases.
3. It is possible that the increase in structuralforces in the lining due to adjacent pile issignificant and cannot be neglected. The
possible increasing structural forces due tofuture construction should be considered intunnel lining design. Otherwise, the lining
must be checked if it can resist the additional
forces due to the planned construction.
ACKNOWLEDGEMENTS

Fig. 8The authors wish to express their gratitude to theLadkrabang Underground and Tunneling Innovation
Center (LUTIC), Thailand Research Fund (TRF) and
1D2 Group Co. Ltd. for the financial support to thefirst author through Grant MSD58I0075.

The change of axial force in tunnel lining
due to adjacent loaded pile in soft clay.

121GEOMATE- Bangkok, Nov. 14-16, 2016[10] Kasper T and Mescgke G, “ A 3D finite element
simulation model for TBM tunnelling in soft
ground” , Vol. 28, Int. J. for Numerical andAnalytical Methods in Geomechanics, 2004, pp.
1441-1460.
[11] Kasper T and Mescgke G, “A numerical study ofthe effect of soil and grout material properties
and cover depth in shield tunnelling” , Vol. 33,
Computers and Geotechnics, 2006, pp. 234-247.
[12] Lueprasert P, Jongpradist P, Heama N and
Suwansawat S, “ 3D- FEM of EPB shield tunnel
excavation using shell element and grouting
layer” , Int. Conf. on GEOMAT, 2016.
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Formulation andverification”, in Proc. of PLAXIS symposium on
Beyond 2000 in Computational Geotechnics,
Rotterdam, Netherlands, 1999, pp. 281-296.
[14] Prust RE, Davies J and Hu S,“ Pressure meter
Investigation for Mass Rapid Transit in
Bangkok, Thailand” Transport. Res. Res. J. of
the Transportation Research Board, 2005, pp.
207-217.
[15] Rukdeechuai T, Jongpradist P, Wonglert A and
Kaewsri T, “ Influence of soilmodels on
numerical simulation of geotechnical works in
Bangkok subsoil” , Vol. 20, EIT Research andDevelopment Journal, 2009, pp. 17-28.
[16] Jongpradist P, Kaewsri T, Sawatparnich A,
Suwansawat S, Youwat S, kongkitkul W and
Sunitsakul J, “ Development of tunnelinginfluence zones for adjacent pile foundations by
numerical analyses” , Vol. 34, Tunnelling andUnderground Space Technology, 2013, pp. 96109.
[17] Heama N, Suwansawat S, Lueprasert P and
Jongpradist P, “Study of lining behavior affectedby adjacent loading pile in MRTA Blue Line
Project” , paper on National Concention on Civil
Engineering, 2015.

REFERENCES[1] Schroeder FC, Potts DM and addenbrooke TI,“ The influence of pile group loading on existingtunnels”, Vol. 54, Geotechnique, 2004, pp. 351362.
[2] Abdel M- Meguid, Rowe RK and Lo KY, “ 3D
effects of surface construction over existing
subway tunnels” , Vol. 2, Geomechanics, 2002,
pp. 447-469.
[3] Dolezalova M, “ Tunnel complex unloaded by a
deep excavation” , Vol.
28, Computer
Geotechnics, 2001, pp.469-493.
[4] Lueprasert P, Jongpradist P, Charoenpak K,
Chaipanna P and Suwansawat S, “ Three
dimensional finite element analysis forpreliminary establishment of tunnel influencezone subject to pile loading” , Vol. 9, Maejo Int.
J. of science technology, 2015, pp. 209-223.
[5] Katebi H, Rezaei AH, Hajialilue M- Bonab andTarifard A, “Assessment the influence of groundstratification, tunnel and surface buildings
specifications on shield tunnel lining loads ( byFEM) ” Vol. 49, Tunnelling and UndergroundSpace Technology, 2015, pp. 67-78.
[6] Bakker KJ, “ Structural design of linings forbored tunnel in soft ground” , Vol. 48, Delft
University of technology, 2003.
[7] Suwansawat S, “ Earth Pressure Balance ( EPB)
Shield Tunneling in Bangkok: Ground Response
and Prediction of Surface Settlements Using
Artificial Neural Networks” , Ph. D. Thesis of
Massachusetts Institute of Technology, 2002.
[8] Bobbier PG, Moller SC and Vermeer PA,“ Bending moments and normal forces in tunnel
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[9] Skempton AW, “ Cast in- situ bored piles in
London Clay” , Geotechnique, 2, 1959, pp. 153173.

122

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

LIFE CYCLE ASSESSMENT ON LEAKAGE RISK FROMCONTAINMENT STRUCTURES AT COASTAL LANDFILL SITES
Misato Sekitani1, Shinya Inazumi2, Suttisak Soralump3 and Ken-ichi Shishido4
1Advanced Course of Architecture and Civil Engineering, Akashi College, National Institute of Technology,
Japan; 2Department of Civil Engineering, Akashi College, National Institute of Technology, Japan;
3
Department of Civil Engineering, Kasetsart University, Thailand; 4TOMEC Corporation, Japan
ABSTRACTThe enclosure function of toxic substances that originate in waste is expected, and the evaluation of achieving
the enclosure function has been obtained in previous research on waste landfill sites. However, the initialconstruction cost is the prime factor for the structural selection of the landfill site in the current approach. Toxic
substances are thought to not leak from coastal landfill sites, and most research has not advanced
countermeasures for when the toxic substances do leak out in the maintenance management planning. The risk
concerning the leakage of toxic substances is assumed to be a decision making standard by especially adapting
the risk assessment in the coastal landfill sites. This is especially relevant when the leakage of the toxic
substances is presumed to occur according to deterioration in the side impervious walls and by the use of repair
methods of the coastal landfill sites.
Keywords: Coastal landfill site, Life cycle cost, Risk management, Side impervious wall
INTRODUCTIONASSESSMENT OF WATER LEAKAGE RISKIN COASTAL LANDFILL SITES

The enclosure function of the toxic substances thatoriginates in waste is expected at the coastal landfillsite. In the coastal landfill site, it has been
understood that the leakage risk of the toxic
substances increases by considering deterioration of
the side impervious wall by previous studies [1].
Therefore, it is necessary to conduct an appropriate
maintenance repair to achieve the enclosure
function of the toxic substances from which thecoastal landfill site is requested at the life cycle.
However, an initial construction cost is the prime
factor for the structural selection of the landfill site
in the current state. In other words, the factor fordamage level generating the leakage of the toxic
substances due to deterioration in an impervious
wall is considered as few [2]. The toxic substances
have been thought to be no leakage from the coastal
landfill site, and most researches have not been
advanced of the countermeasures and so on when
the toxic substances leak out in the maintenancemanagement planning [2].
In this research, utilization of the risk evaluationaccording to a concept of asset management on theleakage of toxic substances with deterioration of asteel-made side impervious wall to selection of arepairing method in a coastal landfill site is
proposed.

Fragility function
It proposes a standard exceedance probability as oneof the risk assessment, and this technique refers to
the fragility function [3]. The fragility function is
the one defined as the conditional probability ofoccurrence to the specified suffering scale.
In this research, the concentration and the total
flux of the toxic substances that leaks to the sea area
as indexes that shows the leakage risk in the coastal
landfill site are applied. The probability to exceed
the limit reference set in these indexes is defined asa performance function (X) and a standard
exceedance probability (FX) such as Eqs. (1) and
(2).
X=

CA
Ct

FX = Rrob. ( X ≤ 1)

(1)
(2)

Where, X is the performance function, Ct is theconcentration or the total flux of the toxic
substances that leaks to the sea area after use t year
passes, CA is the limit reference setting in eachindex, and FX is the standard exceedance
probability.
Ct (concentration or total flux of the toxic
substances that leaks to the sea area after use t year
123GEOMATE- Bangkok, Nov. 14-16, 2016

passes) and CA (limit reference in each index) adopt
the lognormal distribution function because of
non-negative conditions. When the concrete
example is shown for the concentration of theleaking toxic substances, FX means that theconcentration of the leaking toxic substances
calculates a probability beyond the closure andabandonment of coastal landfill sites, and it assumes
as C=0.1 in this research. Therefore, because thelognormal distribution function is set to theprobability density distribution, FX is expressed as
Eq. (3). By the above-mentioned calculation, FX
exceeding the specified limit reference value whenuses t year passes is calculated. A similarcalculation is performed in the evaluation periods of
30 years, and FX to the age of service and the index
is obtained.
1

FX = ∫
02*( 1 " ln x − λ X % *,1exp )− $
' -dx
2πζ X
+* 2 # ζ X & .*substances that leak from the coastal landfill site canbe calculated using the standard exceedance
probability. Moreover, it is necessary to adopt the
LIME for converting it into monetary value tofacilitate the comparison with the cost, when thedamage loss in the total flux of the leaking toxic
substances quantifies it [4]. The estimation ofdamage cost on the damage level due to the total
flux of the toxic substances is shown by Eq. (6)．
𝐸𝐸# =

ζ X = ln (1+ v

2
CA

(3)

(4)

) (1+ v )
2
Ct

(6)

Where, Ek is the damage cost related the total flux ofthe toxic substances in damage level k, Jk is the total
flux of toxic substances in damage level k, ei is thecoefficient converted into monetary value by thekind of the toxic substances, i is the kind of the
toxic substances.
Because it is assumed the leakage of the toxic
substances to the sea area in the coastal landfill site,
the coefficient converted into monetary value in its
discharge to an area of the sea is adopted in thisresearch. Moreover, because the target that the toxic
substances influences are to the ecosystem and tothe human body, the damage coefficient calculated
by adding both of them each other is set as acoefficient converted into monetary value.
The expected loss value is indicated by Eq. (7).

The average (λX) and the standard deviation (ζX)
in the performance function (X) can be led such as
Eqs. (4) and (5) by the median value and thecoefficient of variation of Ct and CA previously
introduced.
λ X = ln µCA − ln µCt) 𝐽𝐽# ×𝑒𝑒(

𝑁𝑁𝑁𝑁𝑁𝑁 =(5)#( 𝑃𝑃𝑃𝑃𝑃𝑃𝑃𝑃(𝑘𝑘)×𝐸𝐸#(7)

Where, NEL is the expected loss value relating to
toxic substances in coastal landfill sites, Prob (k) is
the damage occurrence probability in damage levelk, and Ek is the damage cost related to the total fluxof the toxic substances in damage level k.
The leakage risk of toxic substances considering
the uncertainties to degradation of impervious wallin the coastal landfill site is estimated by adopting
the risk assessment in the above-mentioned.

Where, λX is the average value in the performance
function (X), µCA is the median value in the limit
reference set in each index, µCt is the median value
in the concentration or the total flux of toxic
substances leaking into sea area after use t year
passed, ζX is the standard deviation in performance
function (X), uCA is the variation coefficient in thelimit reference set in each index, uCt is the variation
coefficient in the concentration or the total flux of
toxic substances leaking into sea area after use t
year passed.

Application method of leakage risk assessment incoastal landfill sites
In the leakage risk of the toxic substances in the
coastal landfill site based on deterioration inimpervious wall, the probability density distribution
is calculated in each index, and also a standard
exceedance probability.
When the concentration of the toxic substancesthat leaks to the sea area is assumed to be an index,
C=0.1 is set as a limit reference. Moreover, when
the total flux of the toxic substances that leaks to thesea area is assumed to be an index, the damage levelis set as shown in Table 1, and the limit reference is
set at each damage level. The limit reference basedon the closure and abandonment of coastal landfill
sites in the total flux is set as the value obtained by

Loss function
The loss function is a function that showed the
expected loss value corresponding to the damagelevel and the standard deviation of the expected loss
[3]．The calculating the expected loss (R) by Eq. (6)requires the amount of the damage loss (Ci) at thedamage level and the occurrence probability (Pi) of
the damage level. In this research, the total flux ofthe leaking toxic substances is suitable as an index
that calculates the damage level of the coastal
landfill site.
The occurrence probability of damage level k(Prob(k)) based on the total flux in the toxic
124

GEOMATE- Bangkok, Nov. 14-16, 2016using the flow rate becoming C=0.1 that is theconcentration of the toxic substances in seepage
water under the hydraulic conductivity with
1.0×10-6cm/sec in technical guidelines ofimpervious walls. Moreover, the limit reference
based on technical guidelines of impervious walls is
set from the total flux when the value of technicalguidelines of impervious walls (hydraulicconductivity with 1.0×10-6cm/sec of the side
impervious walls and with 1.0×10-5cm/sec of theclay layers) is adopted. In addition, the probability
density distribution is not set, because it is thoughtthat the environmental quality standard value in theclosure and abandonment of coastal landfill sites is
as appropriate as limit reference (CA) set in theabove-mentioned.
Based on the above-mentioned assumption, it
proposes to the coastal landfill site where theleakage risk can be decreased by evaluating theleakage risk of the toxic substances in the coastal
landfill site, and understanding quantitatively the
influence of each analytical condition from theviewpoint of the risk.

Table 1 Loss level at total flux of seepage
pollution materials

Results of risk assessment concerning leakage

Fig. 1 Probability of excessed standard about
seepage pollution concentration

Figure 1 shows a standard exceedance probabilityby the concentration in each model. The standard
exceedance probability by the total flux of theleaking toxic substances is shown in Fig. 2. The
standard
exceedance
probabilitybytheconcentration rapidly increases after the elapse of 13
years in analyzed model with the steel-made side
double impervious wall (see Fig. 1). The standard
exceedance probability of 10% in 50 years is
adopted as an important probability value on safety
in the infrastructure [3]. This is equivalent to thestandard exceedance probability at the elapsed timeof 1 year is equivalent to 0.2105%. The standard
exceedance probability of an analyzed model withthe steel-made side double impervious wall after theelapse of 12 years is equivalent to 0.04%, andexceeds the above index. Also in an index of thetotal flux by leaking toxic substances, as shown in
Fig. 2, when the closure and abandonment of coastal
landfill sites (rank3) by total flux exceeds thestandard exceedance probability of 10% at theelapse of 50 years, the point is equivalent to theelapse of 15 years in analyzed model with thesteel-made side double impervious wall. In otherwords, it is difficult to say that analyzed model withthe steel-made side double impervious wall satisfies
the containment function of the toxic substancesconsidering the deterioration of impervious walls incoastal landfill sites.

Fig. 2 Probability of excessed standard about
total flux of seepage pollution materials
DECISION OF OPTIMAL MAINTENANCE
STRATEGIES BASED ON RISKASSESSMENT CONCERNING LEAKAGE
OutlineIn the coastal landfill site, it has been understoodthat the leakage risk of the toxic substances
increases by considering deterioration of the side
impervious wall. Therefore, it is necessary to
conduct an appropriate maintenance repair toachieve the enclosure function of the toxic
substances from which the coastal landfill site isrequested at the life cycle.
When the maintenance strategy of social
infrastructures is planned, it is necessary to select anappropriate repairing method, time and so on fromthe concept of an asset management. Moreover,whether it is necessary to repair by using the life
125

GEOMATE- Bangkok, Nov. 14-16, 2016

cycle cost (LCC) and the cost benefit ratio including
the repairing cost and the damage cost as a decision
making index concerning the maintenance strategy
when repairing is judged. However, when the
manifestation of the environmental risk is shown, it
is necessary to repair because it is top priority thatthe landfill site achieves the function to enclose thetoxic substances contained in waste.
In this research, deterioration in the side
impervious wall is assumed as a factor that causes toincreases the environmental risk in the coastal
landfill site, and it is proposed as a countermeasuremethod of repairing the side impervious wall. The
leakage risk of the toxic substances is adopted as adecision making index concerning the repair. That is,an appropriate maintenance strategy that considers
the decrease of the leakage risk according to therepair is examined.

Fig. 3 Relation between total flux andrelative concentration

Assumptions
The leakage risk of the toxic substances according
to deterioration in impervious wall is assumed to bea decision making standard in the examination of
the maintenance strategy of the coastal landfill site,
and the judgment index and the repairing method
are examined in this research The assumptions in
the decision of appropriate maintenance strategies isshown below.
(1) As seawall form of coastal disposal sites, toadopt a steel-made side double impervious wall.
The evaluating the cross-sectional viewassumes analyzed model with the steel-madeside double impervious wall (see Fig. 3).
(2) The evaluation period is assumed to be 30
years.
(3) Because assume the water-level difference in
the inside and outside the landfill site to be
200cm,
sothe
relationshipbetweenconcentration distribution and the total flux ofthe leaking toxic substances obtained in theprevious study is used in examination of themaintenance strategy (see Fig. 4).
(4) The concentration, the total flux of the toxic
substances leaks, and the hydraulic conductivityof the side impervious wall are adopted as a
judgment index to repair.
(5) Propose a strategy to repair steel-made side
impervious walls on sea side only as a repairing
method in coastal landfill site, also a strategy torepair steel-made side impervious walls on theboth sides (sea side and landfill site side), in
analyzed model with the steel-made side double
impervious wall. Moreover, the repair of thesteel-made side impervious wall is made only
once.
(6) The hydraulic conductivity of the steel-madeside impervious wall after repairs is assumed to
be a recovery to the hydraulic conductivity at

Fig. 4 Double steel pile model at evaluated risk
Table 2 Proposing plan of maintenance andrepair in this study

Table 3 Point in time when the standard about
decision index

the time of initial. The deterioration curve of
the steel-made side impervious wall after
repairs adopts the curve of the best case where
deterioration progresses least.
(7) The maintenance strategy sets six patterns in
total as shown in Table 2, by set defining point
as the time to repair that exceeded the standard
in each judgment index.
Examination of appropriate maintenance strategy
based on risk assessment
In the case that repairing method is not applied, theresult indicating a relationship between theconcentration and the total flux is assumed to be a126

GEOMATE- Bangkok, Nov. 14-16, 2016“base-plan”. In the base-plan, point over the
standard set in each judgment index is shown in
Table 3. Set point over the standard becomes earlyby Table 3 in order of the concentration, the total
flux, and the hydraulic conductivity. In eachrepairing method, it is assumed that the side
impervious wall is repaired at the elapsed time ofshowing in Table 3.
In each repairing plan, the transition of astandard
exceedance
probability
fortheconcentration of the leaking toxic substances isshown in Fig. 5. The effect of the repair is notconfirmed from showing no differences withbase-plan in plans 2 and 4. At the elapse of 30 years,
plans 1 and 2 are able to suppress a standard
exceedance probability most. It can be said that it isnecessary to adopt the concentration of the leaking
toxic substances as the judgment index for the repair.
When the side impervious wall on both sides isrepaired, the decrease of the leakage risk can beattempted by the difference of the repairing
methods.
In each repairing strategy, the expected loss
value for the leaking toxic substances and thestandard deviation in the expected loss is calculatedby the loss function. In addition, the comparative
results in the effect of the risk reduction in eachrepairing strategy is shown by plotting it in theplane indicating risk-expectation that used in thefinancial engineering field (see Fig. 6). In Fig. 6, thebase-plan is plotted in upper right. On the other
hand, each repairing strategy is plotted in under left.
In the plane indicating risk-expectation of thisresearch, plotted in upper right mean it is high risk
because the effect caused in leaking toxic
substances and differences are large. On the other
hand, it means the expectation and the standard
deviation are controlled, and it can be called alow-risk to be plotted under the left.
The leakage risk was able to be decreased
compared with the base-plan to not repair one, andthe effect of the repair could be confirmed in eachrepairing strategy. It was shown the plans 1 and 2
are able to decrease expectations most in eachrepairing strategy by the viewpoint of the loss
expectation. Moreover, same as a standard
exceedance probability, it can be said that it is amethod to which the method that repairs the side
impervious wall on both sides decreases the loss
expectation most. The difference is not confirmed in
all repairing strategies from the viewpoint ofstandard deviation. This is because, the deterioration
curve of the side impervious wall after it is repaired
is united to the curve of the best case with the latest
progress of deteriorate, it is convergence thedifference concerning the leakage of the toxic
substances.
The concentration of the toxic substances thatleaks to the sea area is the most suitable for a(a) Repairing side impervious work (sea side)(b) Repairing side impervious work (both side)
Fig. 5 Probability of excessed standard about
relation concentration(a) Repairing side impervious work (sea side)(b) Repairing side impervious work (both side)
Fig. 6 Risk-expectation plane about repair plan
Table 4 Repair and cost effect in repair plan

judgment index for the repair. The decrease in awater interception performance of the side
impervious wall is suppressed because the repairing
strategy of which the judgment index is theconcentration of the sea area compared with otherjudgment indices repairs at the early stage. And, it
127GEOMATE- Bangkok, Nov. 14-16, 2016

can be shown to decrease the leakage risk of the
toxic substances by repairing the side impervious
wall on both sides in the repairing method.
Because the repairing cost and the achieved
effect vary by the difference of the repairing method,
an appropriate repairing strategy is selected by using
a net present value (NPV) and the life-cycle cost
(LCC) in asset management. The repairing cost is
not considered because it pays attention only in theeffect of the risk reduction of the repair. It will benecessary to select the repairing method from theviewpoint of cost-effectiveness because it is thoughtthat the investment assigned to the maintenance ofsocial infrastructures in Japan decreases in the
future. The cost-effectiveness in the repairing
strategy of the coastal landfill site is calculated by
Eq. (8) [5].
𝐸𝐸𝐸𝐸𝐸𝐸𝐸𝐸𝐸𝐸𝐸𝐸 𝑜𝑜𝑜𝑜 𝑐𝑐𝑐𝑐𝑐𝑐𝑐𝑐 𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚𝑚 =:; )<= >=?#(@A B(C# :; ): ):D(E CFGC)?@E=
B=H?(B E:C)

considered was performed. The results and thefindings achieved in this research are shown as
follows.
(1) To consider the uncertainties in the prediction
for a future in the coastal landfill site, theapplication of risk assessment was proposed. In
the risk assessment, a “loss function” that is
applied generally in earthquake risk analysis
could show over standard probability, theprobability that index exceed the limited
standard, loss expectation value in leaking toxic
substances, and standard deviation.
(2) To perform an evaluation of water leaking risk,
it was proposed that an important probability
value on safety in social infrastructures is theexceedance probability of 10% during 50 years,
and that exceeds at the elapse of 15 years in ananalyzed model with the steel-made side double
impervious wall was confirmed. However, it isdifficult to say that the steel-made side double
impervious wall is satisfied the containment
function of toxic substances as the structure oflandfill sites considered the deterioration of thewater interception.
(3) The superiority of the strategy to repair when
the concentration of the toxic substances thatleaked to the sea area exceeded the closure andabandonment of coastal landfill sites wasshown. Moreover, it was confirmed that thestrategy to repair only the side impervious wallon the sea area side was suitable when
considering the viewpoint of cost-effectiveness.
It is clear that the leakage risk of the toxic
substances increases because it considers
deterioration in impervious wall in the coastal
landfill site. It is a critical issue to decide thestrategy on design and the maintenance to consider
the leakage risk of the toxic substances at the life
cycle of the coastal landfill site.

(8)

For calculating the cost-effectiveness of eachrepairing strategy, it is defined as the amount of thedecreasing the leakage risk for the toxic substanceson the effect of repairing the coastal landfill site. In
other words, a repairing cost is assumed to be thematerial cost of required water swelling material torepair the joint section of sides impervious wall witha cross section model (horizontal 1m × vertical 13m
× width 5m) . Table 4 shows the effect of the repair
and cost-effectiveness in each repairing strategy
based on the above-mentioned definition. It can besaid that the repairing method that repairs the side
impervious wall on both sides is appropriate
because plan 4 is the largest for the effect of therepair. However, it is shown that plan 1 isappropriate in the viewpoint of cost-effectiveness.
That is, it is interpreted that the effect of the repairof corresponding to the repairing cost is notachieved though the leakage risk can be decreased
by repaired the side impervious wall on both sides.
It is understood that an appropriate repairing
strategy varies by the decision making standard inplanning an appropriate maintenance strategy. In themaintenance of the coastal landfill site, it isnecessary to repair only the side impervious wall on
the sea side under the limited budget constraint
though it is necessary to repair the side impervious
wall on both sides if it locates with top priority
when putting the decrease of the leakage risk.

REFERENCES[1] Ministry of the Environment: Study Results Report
Towards the Costal Waste Landfill Site Closure or
Abolished Apply Manual Developed (Draft), 2008.
[2] Waterfront Vitalization and Environment Research ofCenter: Management Type Waste Landfill Seawall
Design, Construction, and Management Manual(Revised edition), 2008.
[3] Aoto, H. and Yoshikawa, H.: A Study of Fragility
Curve and Seismic Loss Function for Reinforced
Concrete Pier, Japan Concrete Institute, Vol.30, No.3,
pp.43-48, 2008.
[4] Itsubo, T. and Inaba, A.: Life Cycle Environmental
Impact Assessment Method, MARUZEN, 2005.
[5] Miyagawa, T., Yasuda, K., Iwashiro, I., Yokota, H.
and Hattori, A.: Asset Management for Civil
Engineers -From a Concrete Structures Viewpoint-,
Japan Society of Civil Engineers F, Vol.64, No.1,
pp.24-43, 2008.

CONCLUSIONSA quantitative evaluation on the leakage of the toxic
substances in the coastal landfill site where thedecrease in a water interception performance due todeterioration in the side impervious wall had been128

Sixth International Conferenc e on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-4-5 C3051

ANALYTICAL EVALUATION OF THE INFLUENCE OF HOLESFORMED BY PULLING OUT PILE FOUNDATIONS ON
THE MECHANICAL CHARACTERISTICS OF
THE SURROUNDING GROUND
Kazuki Nosho1, Shinya Inazumi2, Suttisak Soralump3, Shuuichi Kuwahara4 and Yusuke Torigoe51Advanced Course of Architecture and Civil Engineering, Akashi College, National Institute of Technology,
Japan; 2Department of Civil Engineering, Akashi College, National Institute of Technology, Japan;
3
Department of Civil Engineering, Kasetsart University, Thailand; 4Marushinn Co., Ltd., Japan;
5
Department of Earth Sciences, Okayama University, Japan
ABSTRACT

Teardowns of social infrastructure, including civil structures, have been increasing in number in recent years
because these structures have aged and their utilization has decreased along with the decrease in population. The
number pile foundations being pulled out is now far greater than that being newly installed. However, after a pilefoundation is pulled out, the mechanical characteristics of the surrounding ground may be affected by the existenceof the resulting hole formed by pulling out. There are no regulations yet on injecting fillers into pull-out holes, andthe influence of filler strength on the surrounding ground is yet to be elucidated. This study considers the influenceof a pull-out hole on the static and dynamic characteristics of the surrounding ground using two-dimensional
dynamic finite-element analysis. The special qualities required by fillers injected into such holes are also clarified.
Keywords: existing pile foundation, pull-out hole, surrounding ground, two-dimensional dynamic finite-element
analysis
as such materials cannot ensure reliable filling or
stable strength, the use of processing soil and cement–
bentonite has increased in recent years. However,there are no regulations yet on the filler that are
injected into pull-out holes, and the influence of filler
strength on the surrounding ground is yet to beelucidated.
This study considers the influence of the pull-out
holes of pile foundations on the static and dynamic
characteristics of the surrounding grounds using twodimensional dynamic finite-element analysis. The
special qualities required by fillers for being injected
into pull-out holes are also clarified in this study.

INTRODUCTION
In Japan, many cities are located on soft ground,and many structures use pile foundations. Therefore,
for land to be reutilised in places where existing
structures are present, it is necessary to remove theexisting piles supporting the structure as well as the
structure itself for construction of a new structure.
Further, existed piles and concrete husk become
industrial waste, be left of these industrial waste in the
ground is a very difficult problem. In addition, thepresence of such piles remaining in the ground is seen
as a hidden defect in land transactions [1].
Accordingly, it can be said that the removal ofexisting piles is essential.
Methods of removing existing piles include thepull-out method and the crushing-removal method [2].
The crushing-removal method suffers from vibration,
noise and environmental problems, and hence, thepull-out method is more widely used. However, thepull-out method also has certain problems; inparticular, a hole is formed when an existing pile ispulled out of the ground, and if this hole is left
unattended, it is possible that the ground surface may
subside as earth and sand collapse into the hole.
Therefore, it is necessary for filler to be injected into
the pull-out hole. Conventionally, mountain sand or
sand recycled from construction is used as filler, as
these materials are simple and inexpensive. However,

SUMMARY OF THE STUDY OF THE
DYNAMIC BEHAVIOUR OF THE PULL-OUT
HOLE AND THE ORIGINAL GROUND
The tasks involved in this study are described inpoints (1) to (5) (as shown in Fig. 1).
(1) Select the study cross section and the input
ground-motion waveform.
(2) Create an analytical model based on the crosssection selected in point (1), and set the meshdivision of the analytical area.
(3) Select an analytical constant. Set the application
configuration model and material parameters in
the initial stress analysis and the dynamic total
129

GEOMATE- Bangkok, Nov. 14-16, 2016

START(1) Select the study cross section andthe input ground-motion waveform(2) Create an analytical model and mesh(3) Select an analytical constant(a) Empty pull-out hole(4) Perform the initial stress analysis.
(5) Perform dynamic total stress analysis

END
Fig. 1 Analytical procedurestress elastic–plastic analysis.
(4) Perform the initial stress analysis. The analysis
technique used is total stress analysis. In this study,
an HD model for the ground material and anelastic model for the hollow portion are applied.
(5) Took over the calculated ground in stress at 4 ,perform dynamic total stress analysis. Again, aswith the initial stress analysis, the HD model forthe ground material and the elastic model for thehollow portion are applied. Enter the seismic
acceleration at the bottom of the analytical model.

(b) Filled pull-out holeFig. 2 Sectional view

DYNAMIC BEHAVIOUR OF THE PULL-OUT
HOLE AND THE ORIGINAL GROUND, THE
ANALYTICAL MODEL AND THE INPUTGROUND-MOTION WAVEFORM(a) Empty pull-out holeAnalytical model
In the analysis, the analytical cross section has
two layers. The upper layer has clay as a soft stratum,
for which the N-value is approximately 4. The lower
layer consists of a strong formation of gravel serving
as a support layer, for which the N-value isapproximately 50. The width of the analytical crosssection is set at 50 m, the thickness of the clay layer
is 18 m, the thickness of the gravel layer is 8 m andthe total depth of the cross section is 26 m. Two pullout holes exist in the model at a spacing of 4 m; thepore diameter is 1 m, depth is 20 m and the depth ofembedment in the gravel layer is 2 m. To improve theaccuracy of the analysis, a finer mesh spacing is usednear the pull-out hole. This finer mesh continues to be
used even when the pull-out holes are filled in order
to examine the behaviour of the filling. As a boundary(b) Filled pull-out holeFig. 3 Analytical model

condition, in the dynamic analysis, the bottom is a
fixed fulcrum and the lateral boundary is the equaldisplacement boundary. When the moving node on
the side of the left side, node of the other side to the130GEOMATE- Bangkok, Nov. 14-16, 2016

γt represents the weight per unit volume of soil; γw
represents the weight per unit volume of water; G0
represents the initial shear stiffness; σ’m represents theinitial average active confining pressure; ν represents
Poisson’s ratio; c represents the adhesive force; φrepresents the internal friction angle; qu represents thecompressive strength; and E represents the elastic
coefficient.

Table 1 HD model parameters
Parameters
(kPa)
(kPa)
ν (-)
c (kPa)φ (°)

Clay layer
27900
90
0.45
25
0

Gravel layer
298485
234
0.40
0
50

Input ground-motion waveform

Table 2 Elastic model parameters
Parametersqu(N/mm2)E (kN/m2)ν (-)

filler 1

Pull-out holes (filler)
filler 2
filler 3

0.1

0.5

1.0

25280
0.48

1264000.462528000.44

In the analysis, the waveform of the El Centro
1940 NS earthquake (which was provided by the
Building Centre of Japan) is exerted on the substrate’s
surface. The maximum acceleration is 341.7 cm/s2.
ANALYTICAL RESULTS AND STUDY OF
THE DYNAMIC BEHAVIOUR OF THE PULLOUT HOLE AND THE ORIGINAL GROUND

Table 3 Element parameters
Parameters(kN/m3)(kN/m3)
Constitutivelaw

Clay layer

Gravellayer

Pull-out
holes(filler)

15

21

15

The results of the dynamic analysis for the three
types of ground are shown as follows.
Results of the dynamic analysis

9.8
HD modelHD modelFigure 4 shows the diagrams of the maximum
displacement for each of the three ground types. In
addition, Fig. 5 shows the X-direction displacement
contour diagrams, and Fig. 6 shows the Y-direction
displacement contour diagrams. Figure 7 shows thetime history view of the X-direction displacement of
the ground centre in the cases where no pull-out holes
are present and where the pull-out holes have been
filled.

Elastic
model

displacement is the same movement as the node on
the side of the left. Therefore, it is possible to express
whether the stratum has spread to the left or right.
Analysis is performed on three types of ground:
one with no pull-out holes, one with empty pull-out
holes and one where the pull-out holes are filled. The
analytical cross-sectional view of the ground is shownin Fig. 2. A similar analytical model is shown in Fig.
3. In the case of filled pull-out holes in Figure 3, theportion surrounded by a red frame is the pull-out holeportion. The parameters of the HD model in the clay
and gravel layers are shown in Table 1. Table 2 shows
the parameters of the elastic model of the pull-out
hole.

Study of the results of the dynamic analysis
It can be seen that the ground wave causes
significant horizontal displacement in all cases when
it is allowed to act on the base surface. From Figure
4, it can be seen that a large ground subsidence occurs
in the vertical direction when the pull-out hole isempty. The maximum subsidence in ground surface
is approximately 60 cm, which is very dangerous.
However, this subsidence does not occur in the casewhere the pull-out holes are filled. Thus, it can be said
that filling the pull-out hole is an effective way toprevent subsidence.
From Fig. 4, it can be seen that, using filler 1, themaximum horizontal displacement is on the left side,
unlike in the other cases. Figure 5 clearly shows thatthe behaviour of the horizontal displacement underfiller 1 is different from that of the other cases.
In the other cases, displacement to the right occurs
rapidly at approximately 5 seconds and displacement
have continued while the displacement remained.
However, if the displacement at 5 seconds is small,

Constitutive law and material parameters
The parameters in the clay and gravel layers usedin the analysis as well as the soil parameters in thepull-out hole are shown in Table 3. In this analysis, a
fluidising processing soil is used as a filler;
experimental values of its properties can be seen in
the literature [3], [4]. In addition, three fillers withdifferent elastic moduli and different Poisson ratios
are analysed in order to examine the effect of filler
strength on the ground. The strengths of the fillers
increase in the order filler 1, filler 2, filler 3.
Parameters used in the analysis is to determine theanamnestic literature reference [5].
131GEOMATE- Bangkok, Nov. 14-16, 2016(a) No pull-out hole(c) Filler 1(b) Empty pull-out hole(d) Filler 2(e) Filler 3

Fig. 4 Displacement diagram (The amount of displacement is three times)(a) No pull-out hole(b) Filler 1(c) Filler 2(d) Filler 3
Fig. 5 X-direction displacement contour diagrams(a) No pull-out hole(b) Filler 1(c) Filler 2(d) Filler 3
Fig. 6 Y-direction displacement contour diagrams132

GEOMATE- Bangkok, Nov. 14-16, 2016(a) No pull-out hole(b) Filler 1(c) Filler 2(d) Filler 3

Fig. 7 Time history view of the X-direction displacement of the ground centre

portion and subsidence occurs on the right side when
fillers 2 and 3 are used. Furthermore, it can be seenthat the displacement increases as the filler strength
increases.
CONCLUSIONSIn this study, two-dimensional dynamic finiteelement analysis was used to elucidate the effect ofpull-out holes of foundation piles on the dynamic
behaviour of the ground.
The results obtained from the analysis are shown
below.
(1) Ground subsidence occurs in the vicinity of a pullout hole when the hole is left empty, particularlyin the area sandwiched between two pull-out holes.
A very large subsidence occurs in a wide range ofground surfaces if ground motion is applied .
(2) When the pull-out holes are filled, groundsubsidence does not occur in the dynamic or static
analyses. For this reason, it is clear that filling
pull-out holes is effective.
(3) When the filling strength is too small relative tothe strength of the original ground, the ground
may behave in a significantly different way fromthe case with no pull-out holes. For this reason, it
is necessary to change the filler strength to suit the
ground conditions.
This study has not taken account of the influenceof different compounding filler materials or different
hole shapes. Therefore, there is a need to investigate
these conditions as the subject of future analysis.

Fig. 8 Graph comparing the horizontal displacements ofthe ground centre line

then the system will soon return to its original
configuration. From this fact, it is believed that the
behaviour of the horizontal displacement of theground varies greatly when the filling material
strength is too small. From Fig. 5, it can be seen thatthe horizontal displacement is greatest in the casewhere filler 3 is used and the displacement isapproximate in the case where no pull-out holes andfiller 2 are used. Figure 8 shows a comparison of thehorizontal displacements of the ground centre line.
From Fig. 8, it can be seen that rapid displacement of
the clay layer bottom occurs in all cases. In addition,
the behaviour of the ground with holes filled by filler
2 is considered to be closest to the behaviour of theground without pull-out holes. In the verticaldirection, bulging occurs on the left side of the filling
133

GEOMATE- Bangkok, Nov. 14-16, 2016

processing soil”, Summaries of Technical Papersof Annual Meeting Architectural Institute ofJapan, pp.591-592, 2014.
[4] Furugaki Y, “Deformation properties of thebackfill material which has a fluidity and selfhardening”,
TowN-value-up
Management
Report, No.37, pp.41-44, 2011.
[5] Ministry of Land, Infrastructure, Transport andTourism (MLIT), “Public Building Construction
Standard Specification (Building Work) 2013
Edition”, MLIT, 2014.

REFERENCES[1] Murakami T, “Leaving the foundation pile,backfilling denied the defects of ground support
force of the part, agent of accountability violation
was also negative case”, RETIO, No.82, pp.166167, 2011.
[2] Marushicn Co., Ltd., “About the PG Methods”,
http://marushinn.jp/about_chacking.html, 2016.
[3] Furugaki Y, Nakasawa F, Utaka Y, Tokuda K,“Study on the Young's modulus of the fluidized

134

Sixth International Conferenc e on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-4-5 C3051

DEVELOPMENT OF SILICA-BASED SOLIDIFICATION MATERIALSMADE FROM INORGANIC SOLID WASTES
Yuuya Nakase1, Shinya Inazumi2, Suttisak Soralump3, Takashi Shinsaka4, Jun-ichi Yamazaki4,
Ryo Hashimoto5, Tomoyuki Mizuta6 and Yoshihiro Nakagishi71

Advanced Course of Architecture and Civil Engineering, Akashi College, National Institute of Technology,
Japan; 2Department of Civil Engineering, Akashi College, National Institute of Technology, Japan;
3
Department of Civil Engineering, Kasetsart University, Thailand; 4Sanshin Corporation, Japan;
5
Doboku Chishitsu Co., Ltd., Japan; 6Nippon Steel & Sumikin Slag Products Co., Ltd., Japan;
7
Amangu. Corporation, Japan
ABSTRACTIn a series of studies, the authors have developed a powdery silica-based solidification material composed of heattreated inorganic solid wastes with a high content of silica products (such as waste glass and waste fly ash) mixed
with alkali aids. This material is most suitable for use with iron or steel slag, such as that which comes from a blast
furnace or a steel-making process. The powdery silica-based solidification material mixed with iron and steel slag
can be expected to exhibit higher mechanical strength and more predominant characteristics than a powdery
cement-based solidification material. In this study, the solidification mechanism for the mixture of the silica-basedsolidification material, blast furnace slag and water is clarified from the viewpoints of chemistry and mineralogy
by conducting X-ray diffraction analysis, X-ray fluorescence spectrometry and inductively coupled plasma
spectroscopy.
Keywords: silica-based solidification material, solidification mechanism, waste glass
INTRODUCTION

A POWDERY SILICA-BASEDSOLIDIFICATION MATERIAL COMPOSEDOF INORGANIC SOLID WASTES AND
BLAST FURNACE SLAGIn Japan, while natural resources are being
steadily depleted, emissions of waste are increasing
year after year. In a series of studies, the authors havedeveloped a new solidification material using waste
glass discharged from glass industries for the purposeof reducing and effectively using waste glass andblast furnace slag to prevent the depletion of natural
resources [1]. However, although the mechanism forcuring the mixture of silica-based solidificationmaterial, blast furnace slag, and water has beenclarified, the solidification mechanism has not. In thisstudy, the solidification mechanism is clarified fromthe viewpoints of chemistry and mineralogy byconducting X-ray diffraction analysis (XRD analysis),
X-ray fluorescence spectrometry (XRF analysis) and
inductively coupled plasma spectroscopy (ICP
analysis). The applicability of ground-improvement
methods, such as mechanical stirring and highpressure injection mixing, to the material is also
clarified through laboratory blended tests with sandyand clayey materials.

Basic composition of a powdery silica-basedsolidification material
The new powdery silica-based solidificationmaterial is composed of heat-treated inorganic solid
wastes with high contents of silica products (such aswaste glass and waste fly ash) mixed with alkali aids
for drying and grinding. The main chemicalcomponents of this material are clarified by XRF
analysis, and the results are shown in Table 1. As
shown in the table, the main chemical components
include silicon dioxide, sodium oxide and carbon
dioxide. Further, the main mineral components of thepowdery silica-based solidification material areclarified by XRD analysis; as shown in Fig. 1, thesecomponents are sodium metasilicate and calcium
carbonate. The clear peak in the figure indicates thatthe powdery material has a crystalline structure.
Therefore, the material has a dense structure, whichis excellent for preventing acid infiltration [2].

Table 1 The results of XRF analysis of the silica-basedsolidification material
Chemical composition

SiO2

Na2O

CO2

CaO

MgO

Al2O3

Contents (%)

41.7

37.8

9.1

5.8

2.5

1.3

Basic composition of blast-furnace slag powder
Blast-furnace slag powder is blast-furnace slag
135

GEOMATE- Bangkok, Nov. 14-16, 2016

450000C S C
S400000C
S

350000Strength (COUNTS)

S

Table 3 The results of ICP analysis of the silica-basedsolidification material

silica-based solidification material
blast furnace slag

extract from NaOH
solution (30 min)S : Na2SiO3
C : CaCO3

300000S

S

250000S

200000150000S

C
S

100000

S

S
S

S

50000

extract from water extract from water(30 min)(7days)

Si
Na

m
17000

m
4800

m
3900

Ca
Mg

2800
2600

27000
33000

3500
2500

Al
Ti

2100
71

1500
1200

390
70

Remarks"m" cannot measure because of eluting much more, unit (ppm)

0
10

20

30

40

50
2θ (deg)

60

70

80

90

Fig. 1 The results of XRD analysis of the silica-basedsolidification material

Table 4 The results of ICP analysis of blast-furnace slag
extract from NaOH
solution (30 min)Table 2 The results of XRF analysis of the cement-basedsolidification material
Chemical composition
Contents (%)

CaO
42.0

SiO2
29.6

Al2O3
13.8

MgO
5.2

CO2
5.2

SO3
2.0

with a specific surface area that has been adjusted tobe more than 4,000 or 6,000 cm2/g. It is used inconcrete admixtures for its high-strength, high-flow
and low-heat-build-up properties. Powder with a
specific surface area of 3,000 to 4,000 cm2/g is used
as a material in general blast-furnace cement.
Japanese Industrial Standards (JIS) has manufactured
concrete blast-furnace slag powders with specificsurface areas of 3,000, 4,000, 6,000 and 8,000 cm2/g[3]. We clarify the main chemical components of thispowder by XRF analysis and find that they include
calcium oxide, silicon dioxide, aluminium oxide andmagnesium oxide (see Table 2). Furthermore, Fig. 1
shows that blast-furnace slag powder lacks the clear
peaks of the powdery silica-based solidificationmaterial. Because a low-strength broad peak is found,the powder is suggested to be unstable andamorphous, with good reactivity.

extract from water extract from water(30 min)(7days)

Si

4300

1800

7800

Na
Ca

16000
21000

27
13000

27
31000

Mg
Remarks

210

3200
unit (ppm)990

amorphous blast-furnace slag powder is stimulated
and activated. Furthermore, calcium hydroxide madefrom the blast-furnace slag powder participates in a
hydration reaction with silica in high-alkaline
surroundings to create calcium silicate hydrate (C-SH) [4]. It is believed that filling voids in soil structureswith calcium silicate hydrate condenses and cures thesoil structures. Furthermore, the soil surface isactivated by alkaline ions created during the initialstage of the reaction, and Si eluted from the silicabased solidification material creates a silica colloid. It
is considered that silica colloid fills the voids in soil
structures to adhere to activated soil surface, and cure.
It is further supposed that the long-term strength of
the silica-based solidification material is due to a
pozzolanic reaction.
TEST DESCRIPTIONSOLIDIFICATION MECHANISM OF THE
MIXTURE OF THE SILICA-BASEDSOLIDIFICATION MATERIAL AND BLASTFURNACE SLAG

Sample soil and mixing condition
A list of test cases is shown in Table 5. The
mixture of the silica-based solidification material andblast-furnace slag is used as a solidification material.
The silica-based solidification material is white andpowdery and has a SiO2 content of approximately
50%, a density of 2.56 g/cm3 and a grain size of 20 to100 µm.
The slag consists of blast-furnace slag powder.
The compounding ratios of a silica-basedsolidification material (H) and blast-furnace slag
powder (SL) are SL/H = 5, 10 and 15 by weight. For
comparison, a powdery cement-based solidificationmaterial (C) used in high-pressure injection mixing

From the results of the ICP analysis of the silicabased solidification material, as presented in Table 3,
the elution of large quantities of Si, Na and Ca isobserved. ICP analysis of blast-furnace slag powder,as presented in Table 4, indicates the elution of large
quantities of Ca. In the initial stage of the reaction, theeluted Na and Ca atoms become ionised and combinewith hydroxide ions to form sodium hydroxide andcalcium hydroxide. Because of the creation of sodiumhydroxide, the liquid phase shows high pH, and the136GEOMATE- Bangkok, Nov. 14-16, 2016

The testing method
In order to confirm the strength characteristics ofimproved ground, a compression test is conducted.
Test pieces are created using a mould of diameter 50
mm and height 100 mm. The uniaxial compression
test is conducted after the test pieces are cured for 1,
3, 7, 28 and 56 days.
THE RESULTS OF THE TEST

Table 5 List of test cases
No.

solidification material
SL/H
10
15

C

x
x
x
x
x
x
x
x
x
x

14

compounding conditions
ground material
S/M
cray
sandy
2.0
1.5
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x

cement-based sandy
cement-based craysilica-based (SL/H=10) sandy
silica-based (SL/H=10) cray1210

S/M=2.0
8
6
4
2
0
0

7

14

21

28
35
Age (days)

42

49

56

Fig. 2 Change of compressive strength as a function of timefor each solidification material

Differences between soil materials

14

Compressive strength (MN/m2)

The relationship between compressive strengthand age for each solidification material is shown in
Fig. 2. In the case of the powdery cement-basedsolidification material, the sandy material case has ahigher strength than the clayey material case. Thisbehaviour accords with what has been observed in theliterature [5]. On the other hand, in the case of thesilica-based solidification material, clayey materialcases have higher strengths than sandy material cases
after approximately 20 days. Because the grain sizeof Tochi-cray is very small, its specific surface area
is large. Furthermore, Tochi-cray contains soluble
SiO2 and Al2O3 eluted from the clay mineral.
Therefore, the adhesion effect of silica is greater inclayey materials than in sandy materials. Because
increased strength also occurs because of a
pozzolanic reaction, Tochi-cray has a higher strengththan sandy materials.

5
x
x

1
2
3
4
5
6
7
8
9
10
11
12

Compressive strength (MN/m2)

methods is also tested. Each solidification material isused with slurry. The compounding ratio of water (W)and a solidification material (M) is W/M = 130%. In
the case of the powdery cement-based solidificationmaterial, M = C. In the case of the mixture of thesilica-based solidification material and blast-furnaceslag, the volume of the solidification material is thetotal volume of each component (M = H + SL).
Sandy and clayey materials are selected for the
test. The sandy material is Toyoura silica sand (ρs =2.367 g/cm3), and its water content, w, is adjusted tow = 10%. The clayey material is Tochi-cray, made in
Tochigi prefecture, (ρs = 2.724 g/cm3, wL = 34.0%, wp= 17.0%, Ip = 17.0), and its water content is adjusted
for over 12 hours to be w = 40%. The compounding
ratios of the soil material (S) and solidificationmaterial (M) are set to S/M = 2.0 and 1.5 by volume.

sandy (SL/H=5)
sandy (SL/H=10)
sandy (SL/H=15)1210cray (SL/H=5)
cray (SL/H=10)
cray (SL/H=15)

Silica-based solidification material (S/M=2.0)

8
6
4
2
0
0

7

14

21
28
35
Age （days）

42

49

56

Fig. 3 Comparison by compounding ratio

and calcium hydroxide than the powdery cementbased solidification material, because the pozzolanic
reaction advances briskly after 28 days, increasing
strength. Furthermore, it is believed that the
pozzolanic reaction would continue to increase
strength after 56 days.

Long-term strength
The relationship between compressive strengthand age under each of the compounding ratios isshown in Fig. 3. In the case of the powdery silicabased solidification material, a strength increase isfound after 28 days. It is believed that the silica-basedsolidification material includes more silica, alumina

Efficiency of the delay of strength development
As shown in Fig. 2, in the case of the powdery
silica-based solidification material, the strength137

GEOMATE- Bangkok, Nov. 14-16, 2016

increases with age even after 28 days, but theincrement is small. On the other hand, strength
development of the silica-based solidificationmaterial is slower than that of the powdery cementbased solidification material, and the test pieces do
not stand without support until 7 or 14 days later.
Considering the solidification mechanism of thepowdery cement-based solidification material,
hydration occurs immediately and strength increases
slowly. However, because the silica-basedsolidification material needs time to develop latent
hydraulic properties of blast-furnace slag and forsilica to adhere between soil particles, the test pieces
do not stand without support until 7 or 14 days after
their creation.

materials, and effect of the improvement is large.
(5) The silica-based solidification material undergoesa brisk pozzolanic reaction due to the inclusion ofmuch silica, alumina and calcium hydroxide. The
increase in strength is large after 28 days, andcontinues even after 56 days.
(6) Strength development of the silica-basedsolidification material is slower than that of thecement-based solidification material, because
there is efficiency of delay, it is possible to apply
to ground improvement method to need overlap.
This study indicates that the silica-basedsolidification material has excellent delay efficiency
but that the delay in strength development is a general
problem. The reaction is promoted by increasing theamount of silica-based solidification material. It may
be possible to advance strength development, but theexact compound involved in early strength
development is unclear, making this a problem for
future research.

CONCLUSIONSIn this study, the authors developed a silica-basedsolidification material and clarified its solidification
mechanism. The following results were obtained.
(1) From XRD analysis, the blast-furnace slag
powder was found to be amorphous and caused bycuring mixture with silica-based solidificationmaterial and a blast furnace slag powder.
(2) ICP analysis showed that, during the initial stageof the reaction, Na and Ca were eluted from thesilica-based solidification material, causing thepH of liquid phase to increase and activating theblast-furnace slag powder. Si becomes silica
colloid, and an adhesion effect occurs between
soil particles.
(3) ICP analysis also showed that large quantities of
Ca were eluted from the blast-furnace slag powder,creating Ca(OH)2 and C-S-H and causing
hardening through areaction with Si.
(4) Because the adhesion of the effect between soil
perticles due to silica and the development oflong-term strength due to pozzolanic reactions,clayey materials have higher strengths than sandy

REFERENCES[1] Amano K, Ito M, Yanagisaw Y, “According toenvironmental load considered reuse of waste
and a by-product”, Study of environmental
system, vol.25, pp.255-259, 1997.
[2] Matumoto K, Yonekura A, Ito H, “Developmentacid resistance mortar by mixing admixture”,
Proceedings of the Japan Concrete Institute,vol.27, No.1, pp.883-885, 2005.
[3] Yamada T, Nishigata T, “Strength characteristics
of improved soil used mixed solidificationmaterial made from magnesium oxide and a blast
furnace slag powder”, Vol.10, No.1, pp.57-65,
2014.
[4] Kasai J, “Unburnable cement”, Inorganic Materials,
Vol.1, No.252, pp.144-149, 1994.
[5] Japan Cement Association, “The ground improvement
manual by the cement-based solidification material[the second edition] ”, pp.17-48, 1994.

138

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

HYDRAULIC MODEL EXPERIMENTS ON DIMENSIONS OFARMOR UNITS FOR THE RESILIENT BREAKWATAERS
Shota Inoue1, Kiyonobu Kasama2, Mitsunari Hirasawa3,
Kouki Zen4, Zentaro Furukawa5 and Yuichi Yahiro6
1,2,5,6
Graduate School of Engineering, Kyushu University, Japan;
3,4Department of Joint Research on Environment and Disaster in Offshore and Poat Areas,
Kyushu University, Japan

ABSTRACTThe Tohoku-Pacific Ocean Earthquake and tsunami gave serious damages to many breakwaters in Tohoku
region, Japan. The major cause of failure is the tsunami wave force. However, a new type of failure is confirmed:
caused by scouring of the harbor side mound by overflow and seepage flow. In fact, the cause of failure of 9 amongthe 22 destroyed breakwaters was scouring of the mound and any effective measurement against that cause has
not been invented. Therefore, in the previous research we developed a new method to calculate the stable weightof a wave dissipating blocks in consideration of the overflow and seepage flow from the view point of thegeotechnical engineering. In this paper, using this method, the most effective shape of the wave dissipating block
will be described. Furthermore, some sort of hydraulic model experiments with Kamaishi bay breakwater as a
subject have been done in order to see the difference of damage in terms of different 3 height ratio (given by height
divided by radius of blocks) shapes of blocks and 2 ways of laying blocks. Thorough this methodology the
optimum proportion of the block which has a hole like annulus was found. The hydraulic model experiments also
revealed that the most effective height ratio was 1/2 out of 1/4, 1/2 and 3/4. In conclusion, it can be reasonable touse this methodology to determine the optimum shape of wave dissipating blocks for much stronger breakwaters
to the future tsunami.
Keywords: Armor Units, Tsunami breakwater, Seepage flow, Overflow

units is discussed and some sort of hydraulic modelexperiments with Kamaishi bay breakwater as a
subject have been done. The model armor units werehollow cylinder shaped because it is pointed out by
previous study that such a shape enables to reduce
seepage force. the hollow cylinder-shaped block is
assumed as armor units as shown in Fig. 1 and the
optimal proportion was examined. Furthermore, thestability of hollow cylindrical armor units wasinvestigated thorough the hydraulic modelexperiments. This result will be applicable to generalcaisson type composite breakwaters because theparameters such as overflow water depth ordimensions of mound can be changed in accordancewith each breakwater.
The goal of this research is proposing the mosteffective and economical shape for armor units which
can improve existing breakwaters in order to prevent
damage from huge tsunamis assumed in the future.

INTRODUCTION
Many breakwaters were damaged by the March
11, 2011 Tohoku Earthquake off the Pacific coast andstudies about breakwater stability at the time oftsunami overflow are needed. The breakwaters weredestroyed or deformed by roughly three causes;
scouring of mound by tsunami overflow and seepage
flow, horizontal force caused by wave and surface
elevation difference between sea and harbor side, andreduction of bearing capacity of mound induced byseepage flow. Among them, scouring of harbor-side
mound by overflow and seepage flow has beenstudied and developed recently [1], [2]. In ourprevious research the equation to calculate stableweight Ws for rubble constituting mound has been
proposed in consideration of overflow and seepage
flow [3]. Stable weight is defined as the minimum
weight which enables a rubble or an armor unit to bestable to a tsunami and armor units which havesmaller stable weight are more resistant to tsunamis
and more economical.
It is assumed that installing armor units which canlet seepage flow out and be stable against overflow at
harbor-side mound is effective in order to prevent thescouring by overflow and seepage flow. In this paper,using our method, the most effective shape of armor

CaissonH

Rubble moundr
R

Fig. 1
139

Image of armor units application

GEOMATE- Bangkok, Nov. 14-16, 2016

STUDY ON THE OPTIMAL SHAPE OF AN
ARMOR UNIT

Sea Side

Equations to estimate seepage failure andcalculate stable weight for armor units

Caisson Height

CaissonHarbor Side

Figure 2 illustrates that the overflow and seepage
flow through the rubble mound strike the harbor-sidearmor units and its detail situation is shown in Fig. 3.
Since seepage flow can cause the seepage failure ofarmor units before the overflow, the judgement of
whether seepage flow occurs should be performed at
first. That failure can occur when seepage force
exceeds the slope vertical weight of armor unit.
Therefore, the critical hydraulic gradient ic was
calculated theoretically based on this mechanism.
This eq. (1) can be compared with measured
hydraulic gradient in order to judge whether seepage
failure occurs or not.

ic 

Seepage Flow

Rubble moundFig. 2： Frictional resistance(1)： Slope gradientFig. 33Force diagram about armor units

Study on the optimal shape of armor units
In this section, the most effective shape for armor
units described in Fig 1. is discussed by using eq. (2).
At first, external diameter R in Fig.1 is assumed tobe constant and void ratio
e is calculated. That armor units are placed on the
harbor-side mound and their two different ways ofplacement are shown respectively in Fig4.

3




1 e
8k v g  f（i） sin （ Gs  1）r cos 
1

G
S



3

around：Wave forceka  sCD u 6
2flow： Submerged weight

GS  1
cosθ
1 e

3

seepage：Seepage force: where e is void ratio, θ is mound gradient and Gs is
specific weight of an armor unit.
Hudson’s equation which can find the stableweight Ws for armor units hit by overflow wasreconsidered because it was developed without theidea of seepage flow. According to Fig. 2 and 3, theeq. (2) which calculates critical weight when harborside armor units start moving because of overflow
and seepage flow (stable weight Ws) was developed
theoretically.

Ws Overflow andbreakwater

(2)
r/2

: where CD is resistance coefficient, u is overflow
velocity plunging into mound, g is gravitational
acceleration, fr is frictional coefficient, s is unit
weight of armor units, i is hydraulic gradient, kv isvolume coefficient and ka is area coefficient. Basedon the Nakamura’s [4] and Sawaraki’s [5] assumption,in this study, tsunami is regarded as in a steady statebecause tsunamis are very long waves. In this paper,
this eq. (2) is used for calculation of stable weight ofarmor units.

3
R
2

R/2

r/2

R

R

Fig. 4

Placement ways (left: square placement,right: triangle placement)

Since void ratio e is the ratio of the volume ofvoids to the volume of solids, each one is calculated
as below.

esquare 

140

4

1



r
1  
R

2

1

(3)

GEOMATE- Bangkok, Nov. 14-16, 2016

etriangle 

2 3

1



r
1  
R

1
2level difference Δh was kept by the balance of inflow
and outflow and the bottom line was regarded as
impermeable.
Figure 6 describes the outline of the experimental
equipment and Table 1 shows the specifications of theexperimental materials and other conditions. The
mound consists of rubble whose diameter is 2 ~ 19mm and weight is 0.004 ~ 0.027 [N/each]. Whenmound was created, compactions by a rammer weredone for each 150 mm layer of rubble. In order tomodel tsunamis, the pumps generated circulating
flow and the gates adjusted the water surface
elevation. The sea-side level was raised 1 mm per 10
seconds while the harbor-side level was kept fixedly.
Video cameras recorded the breakwater and armor
units under tsunami. The results from water pressure
gages and water velocity gages were used to conduct
analysis. Three different types of model armor units
were hollow cylinder-shaped and made by threedifferent flat-washer in order to make the difference
due to height ratio H/R obvious. Table 2 shows theproperty of armor units. With the experimental
limitation, weights of model armor units are notexactly the same each other but set close values. andTable 3 shows the experimental cases. Aperture ratio
O is ratio of void area per unit area. Experiment ineach case was conducted twice.

(4)

Because these two eq. (3) and (4) have little
differences in form, the void ratio for the triangle
placement is discussed in the following paper.
Therefore, the eq. (2) is expressed as below byinserting eq (4).
Ws ,triangle 22 R    r    1       H    R  
2 3i32 gf r Gs  1 
2 
1

(
/
)

r
R
 

(5)

 s CD 3u 6

3

For simplicity, the armor units located on the crestof the mound are examined and the slope gradient 
is 0. kv, ka were calculated with representative lengthR, height H and inner diameter r of armor units. As aresult, because stable weight is inversely proportional
to the square of H, the higher armor blocks are, thesmaller stable weight can be obtained.
Furthermore, eq. (5) was taken the partial
derivative with r regarding outer diameter R as
constant. The inner diameter r which gives theextreme value of W was named r0. Therefore, themost suitable ratio r0/R was obtained as blow.

r0
8 3i
 1
R
 Gs  1Side view

Unit : [mm]Inflow(6)

Outflow

This equation shows that when hydraulic gradientincreases due to the seepage flow the hollow should
be made larger in order to let the seepage flow out.

Top view
Outflow

Inflow

HYDRAULIC MODEL EXPERIMENTS ON
THE SHAPE OF ARMOR UNITS

400

Outline of the model experiments

Fig. 5. Boundary conditions

Kamaishi bay breakwater was selected as themodel of experiments and its scale is 1/100. It is thedeepest breakwater in the world whose depth is 63 m.
The height of caissons is around 20 m and the rangeof rubble-mound foundation is from -60 m to -27 m.
Therefore, the effect of seabed can be ignored andonly those of overflow and seepage flow can be seen.

：Water pressure gauge：Fluid velocity gauge

185

Sea side：Wave gauge

195

Harbor side
50
430

150

Test conditions

Rubble-mound2046

In order to understand how the stability of armor
units depends on those dimensions, the three types ofarmor units and two ways of placement were adopted.
Boundary conditions on this experiment are
shown in fig. 5. As is shown in this figure, the water

Fig. 6. Schematic of experimental devices

141

Unit：mm

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1. Test conditions

CaissonRubblemoundWater level
difference

Kamaishi bay brealwater (1/100)
Height 195mm
Size
×Breadth 185mm
×Depth 190mm
density

2.03g/cm3

dimensions

Upper base 326mm
Lower base 2046mm
Height 430mm
Gradient 1:2

Particle size

2mm-19mm

Saturated
density

1.81g/cm3

𝐷=

The number of damaged armor units
The total number of armor units

160mm, 40mm, 80mm, 120mm,
145mm, 185mm, 210mm

Square placement
正方形配置
Case1-1
開口率0.411(Case1)
Case Aperture
ratio O
0.394(Case3)
12
1
0.411
0.390(Case5)

14

10

3

0.394

8

5

0.390

Case3-2

Case3-1

6

Case5-1
Case1-2420Water level at
harbor side

[%]

×

(6)

D (%)
ratio(%)
Damage
被害率D

Model

moved horizontally, that is judged to be damaged.
Equation (6) was used to calculate damage ratio.

Case5-2
0

50

100

150

水位差
(mm)h (mm)Water level
difference

25mm (from the crest of the mound)Fig. 7 Damage ratio of armor units placed square
16

Triangle placement
三角形配置
14 開口率 0.320(Case2)
Case Aperture ratio O
0.300(Case4)
12
2
0.320

Shape
Density[g/cm3]Height ratio H/RHeight H[mm]
Outer diameter R[mm]
Inner diameter r[mm]
Weight[N]
SquareAperture placement
ratio O
Triangleplacement

D (%)
ratio(%)
Damage
被害率D

Table 2. Specification of armor units

Hollow cylinder
7.66

4

10

1/4

1/2

3/4

7.2

9.6

13.2

30

22

18

6

8

Case2-1

0.295(Case6)
0.300

Case4-1

0.295

Case2-2

6

Case4-2

4

Case6-1
Case6-2

2
0
0

50

100

150

15

10.5

8.5

0.250

0.235

0.186

Fig. 8 Damage ratio of armor units placed triangularly

0.411

0.394

0.390

0.320

0.300

0.295

Regardless of the difference in the ways ofplacement, it is obvious that the larger the height ratio
H/R is, the less the damage ratio can be. Moving
armor units were seen in any case before h is around
120 mm at which overflow occurred. This is because
they were floated by seepage flow through the rubble
mound. After that, each damage ratio jumped andmany armor units placed at the point where overflow
water dives were damaged.

Table 3. Experimental cases
Height ratio
H/R
CASE 1
1/4
CASE 2
CASE 3
1/2
CASE 4
CASE 5
3/4
CASE 6

Ways ofplacement
SquareTriangle
SquareTriangle
SquareTriangle

水位差(mm)h (mm)Water level
difference

Relationship between the result and parameters
As a result of the experiments, it was found that
there were differences in damage ratio. In fig. 7 and
8, it is obvious that larger height ratio H/R makes
damage ratio lower, which is supported by the eq (5).
Moreover, comparing fig 7 with 8, it was also
confirmed that less aperture ratio O (void ratio e)
makes the armor units stable.

Results of the model experiments
Figure 7 and 8 illustrate the damage ratio D ofarmor units for each case with water level difference
between sea and harbor side. When an armor unit is142

GEOMATE- Bangkok, Nov. 14-16, 2016

harbor-side mound is less than 1. This also confirms
the fact that seepage failure can occur before
overflow.

Safety assessment for the seepage failure
Contour graphs illustrating safety factors for the
seepage failure when h is 40, 80 120 140 mm areshown in Fig. 9. The safety factor is defined to beratio of the critical hydraulic gradient ic which can becalculated by eq. (1) to measured hydraulic gradient i
Frame 001  15 Apr 2016 
thorough the model experiments. Seepage failure isassumed to occur when safety factor is less than 1.

Safety assessment for the seepage failure
Figure 10 shows the relationship between thesafety factor for overflow and seepage failure andoverflow depth h in each case. The safety factor isdefined to be ratio of the weight of a model armor unit
W to the stable weight Ws which can be calculated byeq. (2). Overflow and seepage failure is also assumed
to occur when safety factor is less than 1.

700

V2

600

500

Distance from the bottom (mm)

800
V3:

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1 1.1 1.2 1.3 1.4

Harbor side

Sea side3

Frame 001  15 Apr 2016 

factorSafety 安全率2400Distance from the edge (mm)a)
Distance from the bottom (mm)

300
800

700

Legend■：Square placement▲：Triangle placement
Fractions are height ratio

2.5

V3:

800

h = 40 mm

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1 1.1 1.2 1.3 1.4

1000
Harbor
V1side1200

Sea side

▲3/4
■3/4

1.5
1

▲1/2
■1/2
▲1/4
■1/4

0.5

Frame 001  14 Apr 2016 

V2

600

500
800

0

b) h = 80 mm
Distance from the bottom (mm)

V2

V2

300
500
800

1000

1200

V1
Harbor side

Sea side

Frame 001  15 Apr 2016 

400
700

800

300
500

400

300

Distance from the bottom (mm)

V2

V2

400
600

1000
1200
V1
Distance from the edge (mm)h = 120 mm

c)

300
600
800

500
700

30

40

50

60

It is found that the larger the height ratio H/R is,
the more stable against overflow and seepage failure
can be. This is consistent with the theoretical study
and the experimental result stated above. The fact thatexperiments performed with the same height ratio
have almost the same value implies that the stability
against overflow and seepage failure depends onheight ratio rather than difference in ways ofplacement. Some of the armor units whose heightratio is 3/4 were damaged in the experiments,
however, the safety factor is more than 1. This isbecause that overturning or sliding of armor units was
not considered in theoretical equation.
The model armor units had different unit weight
and the one which had height ratio 3/4 had the largest
unit weight. Therefore, experiments with armor units
which have the same unit weight are needed in orderto reveal the effect of weight.

V3: 00 0.2
0.10.40.20.6
0.30.80.4 0.5
0.61.40.71.6
0.81.80.9 21 2.2
1.12.4
1.2 2.6
1.3 2.8
1.4
V3:
1 1.2

800

Frame 001  14 Apr 2016 

400
600

20

Fig. 10 Comparison of the safety factor for theoverflow and seepage failure

Distance from the edge (mm)

300
600
800

500
700

10

Overflowdepth
h (mm)
越流水深
h [mm]

Frame 001  15 Apr 2016 

400
700

0

V3:
1 1.2
V3: 00 0.2
0.10.40.20.6
0.30.80.4 0.5
0.61.40.71.6
0.8 1.8
0.9 21 2.2
1.1 2.4
1.2 2.6
1.3 2.8
1.4

800

1000

1200

V1
Harbor side

Sea side

CONCLUSION
800

1000

1200

V1
Distancefrom the edge (mm)In this paper, the formula for a stable weight of thehollow cylinder shaped armor units has beendeveloped. This has made clear that armor units
should be as high as possible to a constant diameterin order to be stable against tsunamis with smaller
weight. Furthermore, the ratio of optimal outer
diameter to inner diameter against overflow andd) h = 140 mm
Fig. 9800Safety factor
1000 for seepage
1200 failure (Case6-1)
V1

Even though overflow did not occur when h is
80 mm, the safety factor around the crest of the143GEOMATE- Bangkok, Nov. 14-16, 2016

seepage failure has been found by the equation.
Hydraulic model experiments with three different
shapes of model armor units and two ways ofplacement were performed. This resulted in that thearmor units whose height ratio was 3/4 were morestable than the others. It is also found that seepage
failure can happen before overflow. Moreover,analysis with the proposed equation was consistent
with the experimental results. However, because theexperimental cases are not still enough, further
studies are needed to more accurately estimate theoptimal shape of armor units
In conclusion, the proposed equation is effective
to determine the optimal shape of armor units in orderto improve existing breakwaters against the bigger
tsunamis in the future. More detailed studies orexperiments are needed to develop the new
methodology.

blocks made of dredged soil" (research leader:
Kiyonobu KASAMA).
REFERENCES[1] T. Arikawa, et al., “Failure mechanism andresiliency of breakwaters under tsunami”,
Technical note of the Port and Airport ResearchInstitute. Port and Airport Research Institute,Japan, No. 1269, 2013.
[2] S. Maruyama, et al., “Experimental study on
stability of armor units for harbor-side rubble
mound of composite breakwater against
tsunami”, Journal of JSCE, Ser. B3 (Civil
Engineering in the Ocean), Japan Society of Civil
Engineers, Vol. 68, No. 2, I 7-I 12, 2012.
[3] S. Inoue, et al., “A model experiment for the
stable weight of rubble-mound considering theoverflow and seepage flow”, Journal of JSCE,
Ser. B3 (Civil Engineering in the Ocean), Japan
Society of Civil Engineers, Vol. 71, No. 2, 2015.
[4] T. Sawaraki, A. Onisi (1982) “Destruction
mechanism due to the resonance phenomenon on
the riprap breakwater slope”, Coastal
Engineering Papers, Vol.35, 1982, pp. 428-432.

ACKNOWLEDGEMENTS
This study is a part of the achievement ofResearch Grants Basic Research Science (B):research issue 25289149 "Development of new-styletsunami breakwater that utilize disaster prevention144Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EVALUATION OF POROSITY IN BIOGROUTED SAND USINGMICROFOCUS X-RAY CT
Shumpei Mitsuyama1 , Kazunori Nakashima2, Satoru Kawasaki2
1

Graduate School of Engineering, Hokkaido University, Sapporo, Japan
2
Faculty of Engineering, Hokkaido University, Sapporo, Japan

ABSTRACT
Biogrouting is a method employed for ground improvement based on microbially induced calcium carbonate
precipitation. It is commonly believed that biogrout has environmental and economic benefits. However, there
remains the need to clearly understand the internal structure of biogrouted soil. In this study, we use microfocus
X-ray computed tomography (CT) to evaluate the porosity in biogrouted sand. X-ray CT is useful as a nondistractive inspection tool. First, we prepare small specimens using coral sand at different dilution rates of culture
solution. After carrying out a solidification test for 2 wks., we perform an unconfined compressive strength (UCS)
test and measure the porosity of the specimens. Our aim is to investigate the influence of the dilution rate on the
UCS and the porosity of sand specimens. The results show that a lower dilution rate resulted in a lower sandspecimen porosity and an increase in the UCS. We investigate the precipitation that fills a void. Then, weinvestigate the relationship between UCS and the ratio of porosity. There was a negative correlation between UCS
and porosity, which closely agrees with previous research. We confirm the validity of the result, and we determine
the UCS from the porosity.
Keywords: Sand Solidification, X-ray CT, Ureolytic Bacteria
𝐂𝐎(𝐍𝐇𝟐 )𝟐 + 𝟑𝐇𝟐 𝐎 → 𝟐𝐍𝐇𝟒+ + 𝟐𝐎𝐇 − + 𝐂𝐎𝟐 (1)𝐂𝐎𝟐 + 𝐇𝟐 𝐎 → 𝐇𝐂𝐎𝟑− + 𝐇 +(2)

INTRODUCTION
Recently, artificial beachrock obtained bybiogrout solidification through microbially induced
carbonate precipitation (MICP) [1]–[5] has attracted
a great deal of attention. This method can be applied
to ground improvement and coastal protection[6]–[8],and it is believed that it is environmentally friendly
and low cost. In previous studies, precipitation insand specimens was evaluated by measuring theunconfined compressive strength (UCS). However,
the amount of precipitation varies depending on thelocation [9]. Improving the UCS is key to enabling us
to understand the internal quantity of calcium
carbonate. In this study, we determine experimentallythe porosity, which indicates the proportion of poresin the total volume in the specimen. There are several
ways to measure the porosity, including the mercury
intrusion technique and the liquid saturation method[10]. However, these methods result in the
deformation of internal structures. In this study, weevaluate the porosity using microfocus X-ray
computed tomography (CT). X-ray CTs can visualize
the internal structure quantitatively without any
deformation. Therefore, UCS and porosity can bemeasured using the same sample.
As shown in Figure 1, Microbial ureasecatalyzes the hydrolysis of urea into ammonium andcarbon dioxide (Eqs. (1) and (2)).

The carbonate ions react with calcium ions. Then,
calcium carbonate precipitates between sand grains
and form cementing bonds (Eq. (3)).
+−
𝐂𝐚𝟐 + 𝐇𝐂𝐎−
→ 𝐂𝐚𝐂𝐎𝟑 + 𝐇𝟐 O
𝟑 + 𝐎𝐇(3)

The byproduct, which is the ammonium ion, canresult in an increased pH. In addition, the carbonate
ion can increase the pH. The precipitation of calcium
carbonate tends to take place at higher pH values.

Fig. 1

145

Urea hydorolysis reaction.

GEOMATE- Bangkok, Nov. 14-16, 2016

coefficients of X-ray beams. In this study, S = 200
and B = 0. The voxel size is 25 μm ×25 μm × 40
μm. In the experiment, we obtained 80 slice images
using a TOSCANER 31300 μhd (Toshiba IT &Control Systems Co., Ltd.), which is installed at
Hokkaido University, Japan. The applied tube voltage
was 130 kV, and the maximum tube current was 62
μA. In this experiment, we used the cone-beamscanning mode, and the number of pixels was 1024
× 1024. The brighter areas represent lower density
regions, while the darker areas represent higher
density regions. Using the maximum-likelihood
thresholding method [10], we then determined thethreshold.

METHODWe used Pararhodobacter sp., which wasisolated from the beach sand in Sumuide, Nago,
Okinawa, Japan [11]. The grain size is shown in
Figure 2. We obtained 1 g of bacteria cultivated on
ZoBell 2216E plate medium (polypeptone 5.0 g/L,
yeast extract 1.0 g, and FePO4 0.1 g/L with artificial
seawater) and placed it in 100 mL of ZoBell medium
solution. Then, we shook it at 30°C, 160 rpm for 24
h. In the meanwhile, 40 g of coral sand was dried at
110°C for 24h.
Table 1 Composition of cementation solution(solvent: artificial seawater).
Reagent
Nutrient broth
NH4Cl
NaHCO3CO(NH2)2
CaCl2

Table 2 Test conditions for syringe tests.

Content (g / L)
3.00
10.00
2.12
30.00
55.5

Case number
Case 1Case 2
Case 3
Case 4log(𝑦) = 0.978 log(𝑥) + 2.621

Psss mass percentage (%)

We separated the cells by centrifugation at 30°C,
3000 rpm for 10 min. We extracted 10 mL at thebottom and moved it to a 90 mL ZoBell medium. As
indicated in Figure 3, we set the syringes at 30°C in
an incubator. On the first day, we diluted the culture
solution with ZoBell 2216 medium. The test
conditions of the syringe test are shown in Table 2.
Sixteen mL of culture medium solution and 20 mL ofthe cementation solution were added to each syringe.
Then, we injected 20 mL of cementation solution anddrained every 24 h for 2 wks. We measured the pH
and concentration of Ca2+ every 3 d. At the end of theexperiment, we drained all of the solution in thespecimens.
We also measured the needle-penetration
inclination (diameter = 2.5 cm, height = 7 cm). The
UCS was estimated using a needle penetration device(SH-70, Maruto Testing Machine Company, Tokyo,
Japan). Equation (4) describes the relationshipbetween the UCS (y) and the needle penetration
inclination (x) determined from 114 natural rocksamples and 50 improved soils with cement.

Dilution rate
1
2
5
10

100
90
80
70
60
50
40
30
20
10
0
0.01

0.1
1
Grain size (mm)Fig. 2 Grain size distribution of coral sand.

(4)

In addition, we observed samples using a
microfocus X-ray CT. After performing thesolidification test for 2 wks., we also measured theporosity in order to investigate the influence of thedilution rate on the UCS and the porosity of sandspecimens. In the X-ray CT images, the smallest unit
is called a voxel [12]. A voxel has a CT value as given
in Eq. (5).
CT value = S μ + B(5)Fig. 3 Syringe cementation test.

where S is the slope, B is the bias and μ is attenuation

146

10

GEOMATE- Bangkok, Nov. 14-16, 2016

RESULTS
The sand was cemented, as shown in Figure 4,
and as shown in Figure 5, the UCS increase as thedilution rates decrease. One explanation for this isthat the UCS is related to the number of bacteria.
Moreover, the sand in Case 4 was cemented only at
the surface (2-cm thick). For cases 1 and 2, theestimated UCS value tended to be higher than in cases
3 and 4. For case 1, the UCS at the upper edge is a
maximum of 7 MPa. For case 4, the UCS at thebottom edge is 0 MPa. This point was not solidified.
The bottom UCS is markedly low since the amount ofbacteria is sufficiently small. In addition, we
estimated the amount of CaCO3 in the sand based on
the Ca2+ concentration of the drainage at 3, 6, 9, 12,
and 14 d. From Figure 6, the amount of CaCO3
increased for all cases. However, the slope of Case 1
on the graph is highest. The amount of CaCO3 tended
to rise significantly when we used dilution rates of 1
and 2. We also measured the pH at the drainage. As
shown in Figure 7, the pH decreases with time. From
the above, the amount of bacteria decreased with time.
For the duration of the test period, the pH of all
specimens tended to be lower than the pH at the
beginning. Figure 8 shows X-ray CT images of allcases. Further, the porosity values for all cases areshown in Table 3. The results clarified that a lower
dilution rate results in a decreased porosity of the sandspecimen and an increase in the UCS.

Fig. 5 The UCS of each specimen

Fig.6 The amount of CaCO3

Fig. 4 Syringe specimen of Case 1 after 14 d of curing.

Fig.7 Variation in the pH values of drainage.

147

GEOMATE- Bangkok, Nov. 14-16, 2016

Case 1Case 2

Case 3

Case 4

Fig. 5 X-ray CT images at the upper edgr.

Table 3 The porosity based on the syringe test.

Case number

Porosity at the upper edge (%)

Porosity at the bottom edge (%)

without bacteria

42.9

50.6

Case 131.0

32.7

Case 231.8

34.8

Case 331.6

40.0

Case 4

33.0

39.6

148

GEOMATE- Bangkok, Nov. 14-16, 2016

(3) There is a negative correlation between UCS andporosity.

DISCUSSIONAs shown in Figure 9, there is a relationshipbetween the porosity (n) and UCS (qu) (Eq. (6))(correlation coefficient: 0.5956).
qu = -0.6092 n + 23.787 (R2 = 0.5956)

REFERENCES

(6)

[1]

There is a negative correlation between UCS and theporosity. This negative correlation closely agrees
with previous research. Miyagi and Komiya [13]reported the relationship between UCS and porosity(Eq. (7)) (correlation coefficient: 0.291) of natural
sandy limestone in Okinawa, Japan.
qu = -0.7616 n + 30.55 (R2 = 0.291)

[2]

(7)

The porosity and UCS is found to be smaller than thevalues for natural beachrock. From this graph, if all
voids are filled with precipitation, the strength will
increase to 23.8 MPa. In the case where the porosity
is greater than 39.0%, the specimen is not solidified.
For the same porosity, UCS is different, and this isbecause the initial porosity is different. The numberof samples is low, and more samples are therefore
needed.

[3]

[4]

[5]

Estimated UCS（MPa）

18
this research
Miyagi et al

16
14

[6]

12
10
8

[7]

6
4
2
0

[8]

0.0

10.0

20.0

30.0

40.0

50.0

Porosity（%）
Fig. 6 Relation ship between Estimated UCS and Porosity[9]
CONCLUSION
In this study, we cemented sand specimens
using different dilution rates with a culture solution todetermine the effects of various conditions on the
USC and porosity. The main findings of this study are
as follows:(1) The sand specimens were cemented up to 7 MPa
after 14 d using one dilution rate. The dilution rate
is lower and the UCS is higher.
(2) We observed the filling of the pore with
precipitation.

[10]

[11]

149

A. Al Qabany, K. Soga et al., “Factors
Affecting Efficiency of Microbially Induced
Calcite Precipitation,” Jounal Geotech.
geoenvironmental Eng., Vol. 138, No. 8, pp.
992–1001, 2012.
L. A. Van Paassen, R. Ghose et al.,“Quantifying
BiomediatedGroundImprovement by Ureolysis : Large-Scale
Biogrout Experiment,” Jounal Geotech.
geoenvironmental Eng., Vol. 136, No. 12, pp.
1721–1728, 2010.
V. Achal and S. Kawasaki, “Biogrout : A
Novel
Binding
Materialfor
Soil
Improvement and Concrete Repair,” Front.
Microbiol., Vol. 7, No. 3, pp. 9–11, 2016.
M. Li, Q. Fu et al., “Bio-grout based on
microbially induced sand solidification bymeans of asparaginase activity,” Sci. Rep.,
Vol. 5, pp. 1–9, 2015.
J. T. Dejong, M. B. Fritzges et al.,“Microbially Induced Cementation toControl Sand Response to Undrained Shear,”
Jounal Geotech. geoenvironmental Eng., Vol.
132, no. 12, pp. 1381–1392, 2006.
T. Danjo and S. Kawasaki, “Microbially
Induced Sand Cementation Method Using
Pararhodobacter sp Strain SO1, Inspired by
Beachrock Formation Mechanism,” Min.
Mater. Process. Inst. Japan, No. 1, 2016.
T. Danjo and S. Kawasaki, “A Study of The
Formation Mechanism of Beachrock in
Okinawa, Japan: Toward Making artificial
Rock,” Int. J. GEOMATE, Vol. 5, No. 1, pp.
634–639, 2013.
M. nakibul H. Khan, S. Shimazaki et al.,“Coral Sand Solidification Test ThroughMiceobial Calcium Carbonate Precipitation
Using Pararhodobacter sp.,” Int. J.
GEOMATE, Vol. 11, No. 26, pp. 2665–2670,
2016.
V. S. Whiffin, L. a. van Paassen et al.,“Microbial Carbonate Precipitation as a Soil
Improvement Technique,” Geomicrobiol. J.,
Vol. 24, No. 5, pp. 417–423, 2007.
M. Kato, M. Takahashi et al., “Evaluation ofPorosity and Its Variation in Porous Materials
Using
Microfocus
X-ray
Computed
Tomography Considering the Partial Volume
Effect,” Mater. Trans., Vol. 54, No. 9, pp.
1678–1685, 2013.
Takashi Danjo, “Fundamental study on
development of artificial rock learnt from

GEOMATE- Bangkok, Nov. 14-16, 2016[12]

beachrock formation mechanism,” Grad. Sch.
Eng. Dr. Thesis, 2015.
H. Yamanaka, Satoru Kawasaki et al.,“Segmentation of internal structure of twophase materials using micro-focus X-ray CT,”
Japanese Geotech. J., Vol. 6, No. 2, pp. 273–
284, 2011.

150

[13]

N. Miyagi and Y. Komiya, “Effective
porosity and compression strength of Ryukyu
limestone in Pleistocene Epoch in the
Quanternary period,” Univ. Ryuku Acad.
Rep., Vol. 50, pp. 131–135, 2003.

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

UNIAXIAL COMPRESSION TEST OF UNSATURATED MASADOUNDER CONSTANT DEGREE OF SATURATION CONDITION AND
ITS MODELING

1

Yuhei Kurimoto1, Seiji Kobayashi1, Takashi Tsunemoto2 and Feng Zhang2
Department of Scientific and Engineering Simulation, Nagoya Institute of Technology, Japan;
2
Department of Civil Engineering, Nagoya Institute of Technology, Japan

ABSTRACT
Soils in surface ground are usually unsaturated. Proper modeling of the characteristics of unsaturated soil istherefore absolutely needed. However, due to its complicated mechanical behavior, establishing a constitutive
model for unsaturated soil is far more difficult than that for saturated soil. In this research, uniaxial compression
tests were firstly conducted on unsaturated Masado (Decomposed granite) under constant degree of saturation
condition to investigate the mechanical behavior of the soil at different degree of saturation. Then, based on thetest data, the performances of previously proposed elastoplastic constitutive model for unsaturated soil werechecked carefully, taking the Bishop-type skeleton stress and the degree of saturation as the state variables. It isfound that the mechanical behavior of the unsaturated Masado can be properly described by the proposedelastoplastic constitutive model that takes the skeleton stress and the degree of saturation as the state variables.
Keywords: Unsaturated Soil, Degree of Saturation, Constitutive Model, Uniaxial Compression Testmodel, which is proved to be time consuming andneeds advanced testing technique. The main purposeof this research is to find out the fundamental
behaviors of unsaturated soil, especially the
influence of the degree of saturation with laboratory
test, and establish a unified constitutive model forunsaturated/saturated geomaterials. In this paper,uniaxial compression tests were firstly conducted on
unsaturated Masado under constant degree ofsaturation condition and then the performances ofpreviously proposed elastoplastic constitutive model
for unsaturated soil were checked carefully with thetest data.

INTRODUCTION
Soils, especially in surface layer, may exist at
unsaturated state, whose void is occupied with waterand air. Because of the complicated mechanicalbehavior of unsaturated soil, the application of
constitutive models for unsaturated soil in numericalanalysis for practical engineering problem is muchless than those for saturated soil. However, an
unsaturated soil is not a special soil but the soil
whose degree of saturation, a state variable, issmaller than 1.0. It is necessary to establish aconstitutive model that can describe both
unsaturated state and saturated state under any stresscondition so that it can properly evaluate thedeformation and the failure behavior, such as slope
failure, due to increase and decrease of the watercontent of the geomaterials.
Since the pioneering work by Reference [1], inwhich Barcelona Basic Model (BBM) was proposed
and regarded as one of the basic models forunsaturated soil, a number of elastoplastic
constitutive models have been proposed to describe
the mechanical behavior of unsaturated soil [2]-[7].
In recent years, some constitutive models, using theeffective stress (or skeleton stress) and the degree ofsaturation as the independent state variables, havealso been established by Reference [8]-[10]. The
model has been verified by some tests but is not
sufficient enough. In order to make it applicable inpractical engineering, it is needed to give anappropriate unifying evaluating method for theparameters involved in the proposed constitutive

OUTLINE OF ELEMENTARY TESTS ANDANALYSES
Uniaxial Compression Test of Unsaturated
Masado under Constant Degree of Saturation
Condition
Masado, typical decomposed granite that widelydistributed in southwest Japan, was used as the
specimen in uniaxial compression test. Physicalproperties and the grain size distribution curve of
Masado, which has been sieved to the soil particles
less than 2.0 mm, are shown in Table 1 and Fig. 1.
The specimens (60 mm in diameter and 10 mmheight) were compacted in one layer directly into theoedometer ring at target void ratio e = 0.50. Basedon the compaction curve of Masado, as shown in Fig.
2, as a fundamental test condition, the target watercontent was set to w = 9, 12, 15, 18, 21%, in which151

GEOMATE- Bangkok, Nov. 14-16, 2016

the optimum water content is wopt = 15%, and dryside is w = 9, 12%; wet side is w = 18, 21%.

conducted. Fig. 3 shows the test apparatus foruniaxial compression tests. Pore water pressure wascontrolled and measured through a ceramic disc withan air entry value (AEV) of 1500 kPa, installed at
the bottom of the specimen. Pressure/volume
controller (PVC, GDS product), which can control
the pore pressure or the pore water volume
according to the usage, was implemented in theuniaxial compression test in order to be able toconduct the tests under constant degree of saturation
condition. On the other hand, pore air pressure wasapplied using a pneumatic regulator at the top of thespecimen.
In the tests, the net stress (vnet = v - ua, v: total
stress, ua: pore air pressure) was applied to 20 kPa at
the beginning, and then the suction (s = ua - uw, uw:
pore water pressure) was applied to 100 kPa (ua =500 kPa, uw = 400 kPa). After confirming that thedrainage of water from the specimen became stable,
switch on the control unit to keep the degree ofsaturation being constant, and simultaneously start
applying the net stress from 20 kPa to 965 kPa witha rate of 60 kPa/h. The control technique proposed
by Reference [11], which can adjust the drainage ofwater from the specimen by increasing/decreasing
the pore water pressure under constant pore airpressure, was adopted in order to keep the degree ofsaturation being constant. The expression for thecontrolling of the degree of saturation is given as,

Table 1 Physical properties of Masado
Unit
%
%
g/cm3
%

Specific gravity, Gs
Liquid limit, wL
Plasticity index, Ip
Maximum dry density, dOptimum water content, wopt

Value
2.66
NP
NP
1.85
13.7

Percentage passing (%)

100

80

60

40

20

0
0.001

0.01

0.1

1

10

Particle size (mm)Fig. 1

Grain size distribution curve of Masado

dVw - S r init.dVv  0(1)

Dry density d (g/cm3)

2.0

where, dVw is the change in pore water volume,
Sr(init.)dVv is the product of the initial degree ofsaturation and the change in volume of the voids. In
the uniaxial compression test, dVw is obtained fromthe product of the cross-sectional area and thevertical displacement of the specimen. Meanwhile,dVw is adjusted using the PVC to keep the degree ofsaturation.

1.9
Zero air void curve

1.8

1.7

Pneumatic cylinder

1.6
w = 9%

12%

15%

18%

21%

uw

Dual-tube burettePressure/volume controller
uw

1.5
5

10

15

20

25

Water content w (%)
Fig. 2

Dial gauge
Load cell

Guide ringCompaction curve of Masadoua
Pore air pressure gauge

In order to verify the influence of the initialdegree of saturation on the mechanical behavior of
Masado, uniaxial compression tests of theunsaturated Masado under the condition of constantdegree of saturation with pressuring method were

Consolidation ring Specimen Pedestal Pore water pressure gauge

Fig. 3

152

Schematic illustration of the test apparatus
for uniaxial compression tests

GEOMATE- Bangkok, Nov. 14-16, 2016

the test data and the simulated data. It should be
noted that the material parameters for elastoplastic
constitutive model and the parameters of MCC havebeen unified in all simulation cases.

Performanceof
Proposed
Elastoplastic
Constitutive Model for Unsaturated SoilIn the simulations, an elastoplastic constitutive
model proposed by Reference [8], was used intheoretical simulation of the tests. The model isbased on the modified Cam-Clay model [12], inwhich the Bishop-type skeleton stress (" = vnet +Srs) and the degree of saturation are used as the statevariables, can take into consideration wetting-drying
moisture hysteresis of an unsaturated soil. The
material parameters for the constitutive model andthe parameters of moisture characteristic curve(MCC) are listed in Table 2. Here, it is assumed that
normally consolidated line in unsaturated state(N.C.L.S.) is parallel to the normally consolidated
line in saturated state (N.C.L.) but in a higher
position than N.C.L., which means that under thesame mean skeleton stress, the unsaturated soil cankeep higher the void ratio than those of saturated soil.
In other words, compression index  is always
constant with the degree of saturation. The N.C.L.S.
and the C.S.L.S. are given in the following relations
as,


p 
q
,    0 
pr 
p


(2)

C.S .L.S . : e  Γ Sr    lnp q,    Μ 
pr 
p


(3)

Material parameters forconstitutive model

Compression index, 
0.089
Swelling index, 
0.0080
Critical state parameter, Rcs
4.01
Void ratio, N (p' = 98 kPa on N.C.L.)
0.69
Poisson’s ratio, 
0.25
Parameter of overconsolidation, a
60.0
Parameter of suction, b
20.0
Parameter of overconsolidation, 
2.0State variable of
0.040
overconsolidation, eVoid ratio, Nr (p' = 98 kPa on N.C.L.S.)
0.71
s
Saturated degrees of saturation, Sr
0.85
Residual degrees of saturation, Srr
0.27
Parameter corresponding to
5.0
drying AEV (kPa), Sd
Parameter corresponding to
1.0
wetting AEV (kPa), Sw
Initial stiffness of2000
scanning curve (kPa), kspe
Parameter of shape function, c1
0.014
Parameter of shape function, c2
0.060
Parameter of shape function, c3
50.0
Note: N.C.L. is the normally consolidated line atsaturated state; N.C.L.S. is the normally consolidated
line at unsaturated state.
Parameters of moisturecharacteristic curve

N .C.L.S . : e  N Sr    ln

Table 2 Material parameters for constitutive modeland parameters of MCC of Masado

where, N(Sr) and (Sr) are the void ratios at N.C.L.S.
and C.S.L.S. under the reference mean skeleton
stress (pr = 98 kPa) and certain degree of saturation.
p and q are the mean skeleton stress and the second
invariant of deviator skeleton stress tensor. M is thestress ratio at critical state and has the same value
for saturated and unsaturated states. N(Sr) isexpressed as,Observed
Simulated

250

Suction s (kPa)N N sN Sr   N  sr
Sr  Sr
Sr  Srr

300(4)where, Srs is the saturated degrees of saturation and
Srr is the residual degrees of saturation, which havedefinite physical meaning and can be determined bywater retention test easily. Nr is the void ratio at
N.C.L.S. under the reference mean skeleton stresswhen the degree of saturation is at residual dry state.

200
150

Wetting

100
50

Drying

0
0

20

40

60

80

Degree of saturation Sr (%)

RESULTS AND DISCUSSION
Fig. 4 shows the comparison between the test
data and the simulated data of the water retention
test based on the pressuring method. In the figure, it
is very clear that there is a good agreement between

Fig. 4

153

Moisture characteristic curve of Masado

100

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 3

Physical properties at different stages

Target water content
w = 9%
w = 12%
w = 15%w = 18%w = 21%

Initial condition
w0 (%)e0
Sr0 (%)
8.5
0.642
35.2
12.1
0.650
49.3
15.0
0.650
61.7
17.8
0.651
72.8
21.8
0.660
87.8

w = 9%
w = 12%
w = 15%

w = 9%
w = 12%
w = 15%w = 18%w = 21%

w = 9%
w = 12%
w = 15%

(c)

400

0.50

0.60

Void ratio e

Suction s (kPa)

0.55

300

200

0.55
0.55

0.50

0.50

0.45

100

0.45

0.45
0.40

S = 30% 40% 50% 60% 70% 80% 90%
r

500

0

0.40
0

5

10

w = 18%w = 21%

0.65

(b)

(a)
0.60

Stage endef
Srf (%)
0.453
31.1
0.479
41.0
0.446
47.7
0.441
48.8
0.455
49.6

wf (%)
5.3
7.3
8.0
7.9
8.5

w = 18%w = 21%500

0.65

Void ratio eStage beginningw (%)
eSr (%)
7.4
0.613
31.6
9.6
0.625
40.7
11.0
0.625
46.8
11.1
0.613
48.0
11.5
0.623
49.0

0

15

200

400

600

800

1

Vertical net stress vnet (kPa)

Water content w (%)

600

700

800

900

1000

0.40

1000

10

100

1000

10000

Vertical effective stress " (kPa)

Fig. 5 Uniaxial compression tests results under constant degree of saturation; (a) e vs. w, (b) s vs. vnet, (c) e
vs. "
S = 35.0%

S = 35.0%

S = 40.0%

r

r

0.70

500

(b)

(a)

(c)

400

30

20

0.65

Void ratio e

Suction s (kPa)

40

300

200

10

100

0

0

r

S = 37.5%

r50

S = 40.0%rr

S = 37.5%r

Degree of saturation Sr (%)

S = 35.0%

S = 40.0%rr

S = 37.5%0.60
0.55000.54980.55
0.5496

0.54940.50
0.5492

0

200

400

600

800

Vertical net stress vnet (kPa)

1000

0.5490

0

200

400

600

800

Vertical net stress vnet (kPa)

1000

480

0.45
1

490

10

500

100

1000

10000

Vertical effective stress " (kPa)

Fig. 6 Simulated results of uniaxial compression tests under constant degree of saturation; (a) Sr vs. vnet, (b) s
vs. vnet, (c) e vs. "
Fig. 5 shows the test results obtained from theuniaxial compression tests under constant degree ofsaturation. The constant degree of saturation lines
were drawn based on eSr = wGs. See Table 3 for thephysical properties of Masado in each test. Fromthese results, it is known that the water content
declines with the decreasing of the void ratio, while
the degree of saturation is kept constant. In Fig. 5(b),the suction is built up along with the increasing of
the net stress, and the increment of suction isdepending on the initial degree of saturation. In twocases (w = 18, 21%), even the degrees of saturation
at the beginning of suction are almost the same, thespecimen with higher initial degree of saturationshows lower increment of suction. As shown in Fig.
5(c), the N.C.L. of dry side (w = 9, 12%) is locatedabove those of wet side (w = 18, 21%) and the
optimum water content (wopt = 15%). In other words,
the lower the initial degree of saturation is, the more
the N.C.L. will be shifted upward.
Fig. 6 shows the simulated results of uniaxial
compression tests under constant degree ofsaturation. The simulated loading path is the same asthat of the tests. In the simulations, the target
degrees of saturation were set into three cases (Sr =35.0, 37.5, 40.0%), which are corresponded to the154GEOMATE- Bangkok, Nov. 14-16, 2016

unsaturated soils”, Soils and Foundations, Vol.
33, No. 4, Dec. 1993, pp. 49-63.
[3] Cui YJ, Delage P, “Yielding and plastic
behaviour of an unsaturated compacted silt”,
Géotechnique, Vol. 46, No. 2, June 1996, pp.
291-311.
[4] Loret B, Khalili N, “An effective stress elasticplastic model for unsaturated porous media”,
Mechanics of Materials, Vol. 34, No. 2, Feb.
2002, pp. 97-116.
[5] Chiu CF, Ng CWW, “A state-dependent elastoplastic model for saturated and unsaturated
soils”, Géotechnique, Vol. 53, No. 9, Nov. 2003,
pp. 809-829.
[6] Sun DA, Cui HB, Matsuoka H, Sheng DC, “A
three-dimensional elastoplastic model forunsaturated compacted soils with hydraulichysteresis”, Soils and Foundations, Vol. 47, No.
2, Apr. 2007, pp. 253-264.
[7] Sheng D, Fredlund DG, Gens A, “A newmodelling approach for unsaturated soils using
independent stress variables”, Canadian
Geotechnical Journal, Vol. 45, No. 4, Apr. 2008,
pp. 511-534.
[8] Zhang F, Ikariya T, “A new model forunsaturated soil using skeleton stress and degreeof saturation as state variables”, Soils andFoundations, Vol. 51, No. 1, Feb. 2011, pp. 6781.
[9] Zhou AN, Sheng D, Sloan SW, Gens A,“Interpretation of unsaturated soil behaviour inthe stress-saturation space: II: Constitutive
relationships and validations”, Computers andGeotechnics, Vol. 43, June 2012, pp. 111-123.
[10] Zhou AN, Sheng D, Sloan SW, Gens A,“Interpretation of unsaturated soil behaviour inthe stress - Saturation space, I: Volume change
and water retention behaviour”, Computers andGeotechnics, Vol. 43, June 2012, pp. 178-187.
[11] Burton GJ, Pineda JA, Sheng D, Airey DW,
Zhang
F,“Exploring
one-dimensional
compression of compacted clay under constantdegree of saturation paths”, Géotechnique, Vol.
66, No. 5, May 2016, pp. 435-440.
[12] Schofield AN, Wroth CP, Critical State Soil
Mechanics, McGraw-Hill, 1968.

value located in the inner side of MCC from Fig. 4.
From these results, the degree of saturation
decreases due to the suction, after then the degree ofsaturation is kept constant by the increment of the
net stress. In Fig. 6(b), it can be observed that thesimulated data is on the whole agreed with the test
data. As shown in Fig. 6(c), the proposed model canqualititatively describe the parallel upward movingof N.C.L. along with the decreasing of the initialdegree of saturation, though quantitatively there still
exists some discrepancies.
CONCLUSIONSIn this paper, uniaxial compression tests wereconducted to verify the importance of proper
selection of the state variables, that is, the skeleton
stress and the degree of saturation as the statevariables, in proposing constitutive model forunsaturated
soil.
Corresponding
theoreticalsimulation were also conducted with proper MCC.
The following conclusions can be drawn:
1. By using the skeleton stress as a state variable,
the influence of the degree of saturation on the
deformation and the strength of unsaturated
soil can be identified.
2. In the simulations, it is proved that the
proposed model can describe the mechanicalbehaviors of unsaturated soil under different
degrees of saturation to some extent on thewhole. However, the comparison between thetest data and the simulated data has not beenchecked directly yet, because the range of thedegree of saturation depending on the suction
can be applied for the simulation is limited. In
future study, it is necessary to determine theparameters, especially for MCC in a moreaccurate way, based on more test cases.
REFERENCES[1] Alonso EE, Gens A, Josa A, “A constitutive
model
for
partiallysaturatedsoils”,
Géotechnique, Vol. 40, No. 3, Sep. 1990, pp.
405-430.
[2] Kohgo Y, Nakano M, Miyazaki T, “Theoreticalaspects of constitutive modelling for155Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

AN ANALYTICAL EVALUATION OF RADIAL CONSOLIDATIONWITH RESPECT TO DRAIN DEGRADATION
1,2,3

Thanh Trung Nguyen1, Buddhima Indraratna2 and Cholachat Rujikiatkamjorn3
Faculty of Engineering and Information Sciences, University of Wollongong, Australia
ABSTRACTThe use of synthetic prefabricated vertical drains over the past decades for soil improvement has had anegative ecological effect on the natural environment due to their long-term existence in soil. The emergence inrecent years of Natural Fibre Drains (NFDs) made from biodegradable materials such as jute, coir, and straw isbelieved to be an environmentally friendly approach to treating soft soil. However, when subjected to adverse
conditions such as pyritic acidic and/or biologically active soil, natural fibres can decay rapidly, and that results
in a significant deterioration of the engineering characteristics of drains during consolidation. This deterioration
can particularly reduce the discharge capacity of the drains, which retards the dissipation of excess pore pressureand delays consolidation. This paper represents an analytical method which can capture the effect that drain
degradation has on the radial consolidation of soil over time, and then applies this method to an exponential
reduction of drain discharge capacity. This analytical method, with supporting evidence from experimental data,
reveals that there is a significant retardation in excess pore pressure dissipation due to drain degradation. The
study indicates there is a potential risk when using NFDs in adverse soils and suggests that caution is needed inpractice.

Keywords: Soil Consolidation; Discharge Capacity; Natural Fibre Drains; Prefabricated Vertical Drains;
Biodegradable Materials

INTRODUCTION
Prefabricated vertical drains (PVDs) have beenused worldwide to accelerate soil consolidation, andover the years have undergone extensive technicalimprovements such as combining vertical drains
with surcharge and vacuum loading [1]. Rather thanusing polymeric PVDs which can damage thenatural environment, natural fibre drains (NFDs)
have been widely used in South and Southeast Asia
in recent years, ever since the first NFDs wereintroduced by Lee et al. [2] in 1987 while asignificant effort [3-5] has been made toconvenience the application of those naturally
occurring materials in geoengineering. Most fieldobservations [4, 6] indicated that NFDs haveengineering characteristics that are comparable tosynthetic prefabricated vertical drains and hence
accelerate consolidation very affectively. Howeverthese NFDs had been used in more inert soft soils
which might not harm the engineering features ofthe drain seriously, but in highly acidic and/or bioactive soils such as acidic sulphate soil where thesulphate reducing bacteria consume organic
compounds for their metabolism, natural fibres
potentially decay much faster, and therefore thedegradation of NFDs in adverse soils needs urgent
evaluation.
Natural fibre drains can be made fromagriculture products such as jute, coir, and straw

with either circular or band shaped cross-sections.
Of these, jute and coir are preferred in practices dueto their favorable engineering characteristics andsheer abundance in developing countries. Coir fibre
extracted from coconut husks has approximately
30% lignin [7] making it more robust and durable
than other natural fibres such as jute which contains
more than 80% cellulose and only around 12%
lignin [8]. Straw has also been employed in several
studies [4] but the drain made from this fibre showed
relatively smaller discharge capacity than thosecomposed of jute and coir.
The degradation of jute fibre basically depends
on the environmental conditions, particularly theacidity and biological activities of soil. For example,
jute fibre can degrade much faster where the pH
value is less than 5.2 or higher than 9 [8, 9]. Since
acidic soil can be found in many regions around theworld, as reported by [10, 11], the effect of this soilon the performance of NFDs must be considered.
Moreover, bacteria and fungi in their preferredenvironment can also enhance the biodegradation ofnatural fibres, as shown by [4, 9]. Several anaerobic
bacteria can decompose organic matters faster when
they have access to oxidants such as 𝐹𝑒 3+ , 𝑀𝑛2+
and 𝑆𝑂42− [12].
This paper discusses the effect that drain
degradation has on the radial consolidation of soil,
based on an analytical approach. The result from this156

GEOMATE- Bangkok, Nov. 14-16, 2016

analytical method is then compared with thoseobtained from an experimental investigation.

which the jute fibres were converted completely into
organic components of soil.

ANALYTICAL APPROACH FOR RADIALCONSOLIDATION WITH RESPECT TO
DRAIN DEGRADATION

Incorporating Drain Degradation into the Radial
Consolidation of Soil
Based on numerous studies into the consolidationof soil, the radial consolidation theory proposed byBarron [13] and later improved by Hansbo [14] isthe preferred approach ([15, 16]). According to thismethod, the dissipation of EPP is described by:Degradation of natural fibre drains
Since the major function of NFD is to discharge
the excess pore pressure (EPP) of soil, theengineering characteristics of the drain are expected
to remain until the design target of consolidation isachieved (Fig. 1). However previous studies ([4, 8,9]) indicated that natural fibres (i.e., jute, straw) can
decay rapidly in adverse environments, leading to aconsiderable retardation in the dissipation of EPP.
Therefore the over decay of NFDs on consolidationneeds serious evaluation.

Drainage characteristics

Service period

𝑢(𝑡)
−8𝑇ℎ
= 𝑒𝑥𝑝⁡()
𝑢𝑜𝜇

In the above, Th is the time factor for horizontal
drainage: 𝑇ℎ = (𝑐ℎ 𝑡)/(𝑑𝑒2 ) where ch is theconsolidation coefficient for horizontal drainage,
and de is the diameter of the influence zone. uo is theinitial EPP. 𝜇 = 𝜇𝑞 + ⁡ 𝜇𝑛,𝑠 where 𝜇𝑞 = (2𝜋𝑘ℎ 𝑙 2 )/(3𝑞𝑤 ) represents the influence of the discharge
capacity qw of a drain with a length l; and n,sdenotes the effect of geometry (i.e., the size of thesmear and influence zones), and is estimated asfollows:

Design target

Normal degradation

Natural fibres
turned totally
into organic
content of soilqw(t)Serious degradation

𝜇𝑛,𝑠 =

Time

Consolidation degree
(%)

0

100

EPP retarded(1)

Seriousdegradation

Normal case

𝑛2
𝑛
𝑘ℎ
3
[𝑙𝑛 ( ) + 𝑙𝑛(𝑠) − ]
2
𝑛 −1
𝑠
𝑘𝑠
4
𝑠2
𝑠2
+ 2
[1 − 2 ]
𝑛 −1
4𝑛
𝑘ℎ
1
𝑠4 − 1+ ( 2)[
𝑘𝑠 𝑛 − 1
4𝑛2(2)− 𝑠 2 + 1]

Fig. 1 Degradation of NFDs in relation to
consolidation process of soil

In the above, n = re/rw; s = rs/rw where re, rs and rw
are the radius of the influence, smear and well zones,
respectively; kh and ks are the coefficient of
permeability in the undisturbed and smear zones,
respectively.
By assuming that the geometric parameters of
NFD are constant and only the discharge capacity ofthe drain decreases over time when natural fibres aredecaying, a time-dependent function of discharge
capacity qw(t) is introduced to the conventional
method of Hansbo. According to Indraratna et al.
[15], the average excess pore pressure of a unit cellat time t can be written by:

The biodegradation of NFDs is complicated anddepends on a number of factors such as: (i) theenvironmental conditions of soil such as pH,
temperature, activities of micro-organisms and soon; (ii) the properties of natural fibres in the draine.g., their chemical components. A pilot
investigation in the field carried out by Kim and Cho[6] on the reduction in the discharge capacity ofseveral natural fibre drains made from straw and juteinstalled in saturated clay indicated that the rate ofdegradation varied considerably over different
seasons where the temperature and humidity
changed.
For simplicity in this study, a number ofassumptions were made: (i) the NFDs began todegrade immediately after installation; (ii) thegeometric parameters (i.e., equivalent diameter andlength) of drains were constant while the discharge
capacity varied over time; (iii) degradation was
uniform over the installation depth; and (iv) the
reduced discharge capacity reached a limit level in

𝑢=
𝑙 𝑟𝑠
𝑙 𝑟𝑒
∫0 ∫𝑟 𝑢𝑠 (2𝜋𝑟)𝑑𝑟𝑑𝑧 + ∫0 ∫𝑟 𝑢𝑢 (2𝜋𝑟)𝑑𝑟𝑑𝑧
𝑤
𝑠
=
𝑉

(3)

In the above, V is the total volume of the unit cell.
By integrating Eq. [3] with the constant geometricparameters and varying the discharge capacity overtime, and then re-arranging yield:
157

GEOMATE- Bangkok, Nov. 14-16, 2016

𝑑𝑢𝑢 + ℎ(𝑡)=0𝑑𝑡discharge capacity of a drain. Note that when approaches zero (no degradation of drain), Eq. [10]turns into Hansbo’s solution which assumes aconstant discharge capacity.

(4)

where h(t) is written as:
𝜆ℎ(𝑡) = 𝜒 (𝜇𝑛,𝑠 +)𝑞𝑤 (𝑡)

Results and discussion(5)

A prefabricated vertical jute drain (PVJD) withan equivalent diameter dw of 70 mm was used todemonstrate the proposed model with respect to theapproximation 𝑑𝑤 = 2(𝑎𝑤 + 𝑏𝑤 )/𝜋 where aw andbw are the thickness and width of the drain. A typical
PVJD is normally 80-110 mm wide and 6-12 mm
thick (Fig. 2). The drain had an initial discharge
capacity qwo of 0.43 m3/day determined by carryingout the discharge capacity test in the laboratory.
Note that PVJDs can be installed using the same
method as conventional drains, so the smear
behaviour of soil using PVJDs should be identical toconventional cases already clarified by Indraratna
and Redana [20]. In this investigation, a ratio of ks/kh= 3 was used, and the soil properties were assumed.

where
𝜆 =⁡

2𝜋𝑘ℎ 𝑙 2
3

(6)

𝜒 =⁡

𝑑𝑒2
8𝑐ℎ

(7)

Solving the ordinary differential equation Eq. [4]and re-arranging result in:
𝑡

𝑢(𝑡)
1
= 𝑒𝑥𝑝 (− ∫
𝑑𝑡)
𝑢𝑜
ℎ(𝑡)

(8)

80-110 mm
6-12 mm0Note that this equation describes the radialconsolidation of soil with respect to any given
degradation function of the discharge capacity qw(t)incorporated into Eq. [4]. In order to demonstrate theproposed solution, a specific form of qw(t) is
required. By referring to biological studies [17-19]
which used mathematical models to predict an
exponential decay of organic matters (e.g., jute,straw), this paper assumes an exponential reductionin the discharge capacity over time, particularly:
𝑞𝑤 (𝑡) = 𝑞𝑤𝑜 𝑒 −𝜔𝑡

𝑢(𝑡)
𝑢𝑜

𝜇𝑛,𝑠
+ 𝑒 𝜔𝑡𝜇𝑞𝑜1+[𝑙𝑛 ( 𝜇)]}𝑛,𝑠
𝜒𝜇𝑛,𝑠 𝜔+1𝜇𝑞𝑜coir fibre
bundleFig. 3 shows how the proposed analyticalsolution could capture the retardation of thedissipation of EPP due to drain degradationcompared to the conventional method without
degradation. In this comparison, the exponential
degradation of the drain discharge capacity with  =0.02 day-1 was considered. In the period where thedrain discharge capacity was large enough (initially
50 days), degradation had almost no influence on thedissipation of EPP, so deviation between the two
curves is not clear, but after this period when the
discharge capacity of the drain was less than 0.1
m3/day, dissipation of EPP induced by thedegradable drain definitely deviated from theconventional drain. The proposed solution indicatedthere was a serious obstruction of EPP discharged
after approximately 200 days meanwhile this trend
could not be shown in the conventional solution
which did not consider any drain degradation. This
result also showed a critical state of decay where thepredicted curve was almost horizontal after nearly

Replacing Eq. [9] into h(t) of Eq. [8] andintegrating lead to:

−8𝑇ℎ
𝜇𝑛,𝑠

Stitches

Fig. 2 Structure of a typical prefabricated verticaljute drain (PVJD)(9)where  is the decay coefficient that represents therate at which the drain discharge capacity degrades,and qwo is the initial discharge capacity of the drain.
Apparently  should be within the range [0, +].

= 𝑒𝑥𝑝⁡{

Jute burlap(10)

The above describes the dissipation of EPP overtime with respect to an exponential reduction in the158

GEOMATE- Bangkok, Nov. 14-16, 2016

300 days at a residual EPP of almost 20%. In thiscritical state the discharge capacity of the drainreached an extremely low value (i.e., below 0.01
m3/day), leading to a very low rate of dissipation of
EPP.

kh = 2.5x10-9 m/s;
ch = 1.58 m2/year
de = 0.7 m; kh/ks = 3
s = 3.4; l = 20m
kh = 2.5x10-9 m/s;
ch = 1.58 m2/year
de = 0.7 m; kh/ks = 3
s = 3.4; l = 20m

Fig. 4 Degradation with different decay rates: a)
Discharge capacity; and b) Dissipation of EPP
Fig. 5 compares the consolidation curves
obtained by the proposed analytical solution and theexperimental work conducted in the laboratory by
Kim et al. [21]. A unit cell with de = 0.6 m; ds = 0.3m; dw = 0.05 m and l = 2 m was adopted forcomputation with reference to the laboratory datameasured by Kim et al. [21]. The drain with an
initial discharge capacity of qwo = 0.014 m3/day anda degradation coefficient of  = 0.260 day-1 (Fig. 5a)
was also obtained from the work by Kim et al. [21].
The soil parameters, including kh = 3.6x10-10 m/s andch = 3.154 m2/year, and the ratio kh/ks = 1.05 wereused in this study.
Fig. 5b shows a good agreement between theanalytical method and the laboratory data measured
by Kim et al. [21]. In the first 7 days, the analyticalcurve was almost 5% higher than the laboratory
curve, but it gradually became closer to theexperimental curve over the next 10 days before
turning into a smaller range. This deviation isunderstandable because of a difference in the draindischarge capacity between the measured and fitted
curves, as shown in Fig. 5a. With  = 0.260 day-1 ofthe exponential degradation adopted, the sharp dropin discharge capacity in the first days of theexperiment could not be captured accurately,
resulting in a gap between the two curves, as shownin Fig. 5b. When the degradation of drain discharge
capacity observed in the experiment was described
better by the exponential curve, the dissipation of
EPP was captured more accurately by the proposed
method.

Fig. 3 Degradation versus no-degradation of drain:a) Discharge capacity; and b) Dissipation of EPP
The radial consolidation of soil induced by adegradable drain with respect to various rates ( =0.01; 0.02; and 0.03 day-1) of degradation isrepresented in Fig. 4. The fastest drain decay
occurred ( = 0.03 day-1) where the drain discharge
capacity decreased to 7.0x10-3 m3/s in 150 days; here
the dissipation of EPP began slow after almost 60
days and then turned into a critical state of decay
after only 160 days with a residual EPP of 35%. In
the smaller range of degradation rates i.e.,  = 0.02
and 0.01 day-1, the consolidation curves reached acritical state after 200 days and 400 days with aresidual EPP of approximately 20% and 5%,
respectively. Unlike the case where the drain did not
decay, the degradable drain caused an obvious
retardation in the consolidation process.
This study also showed that the negative
influence of drain degradation only became obvious
when the discharge capacity decreased to a certain
small value which varies with the properties of soil
(e.g., ch) and geometric parameters such as thelength and space of the drain. In this investigation,
the discharge capacity that began to markedly
obstruct the dissipation of EPP was approximately
0.08 m3/day.

159

GEOMATE- Bangkok, Nov. 14-16, 2016

ACKNOWLEDGEMENTS
The authors acknowledge the Australia ResearchCouncil for funding this research. The 1st author’s
PhD scholarship is sponsored by the Australia
Endeavour Scheme.
REFERENCES

Fig. 5 Model verification by comparing withexperimental work: a) Exponential curve fitting
experimental data; b) Dissipation of EPP
This investigation also indicated that the form of
the degradation function qw(t) played a decisive rolein accurately predicting the influence of drain
degradation on the consolidation of soil. In thispaper, only the exponential degradation of the drain
was considered but it is noteworthy that the
proposed analytical solution (Eq. 8) can be applied
to any given form of qw(t), such as convex andconcave curves. To determine qw(t), it requires anextensive investigation into the degradation of NFDs
subjected to specific environmental conditions ofsoil.
CONCLUSION
An analytical approach which incorporated thedegradation of natural fibre drain into the radialconsolidation of soil was represented in this study,
and then the solution was applied to an exponential
degradation of drain discharge capacity. The results
showed that the dissipation of excess pore pressure(EPP) could be retarded significantly when the
discharge capacity of the drain decreased to less than
0.08 m3/day. Consolidation of soil could be slowed
down seriously when the decay of a drain turned intoa critical state where the discharge capacity becamevery small. This study also indicated there is a needto determine the exact form of the degradation
function qw(t) with respect to specific soil conditions
in order to optimise the accuracy of this method.

[1] Indraratna B, Rujikiatkamjorn C, Kelly C,
Buys H. Soft soil foundation improved by
vacuum and surcharge loading. Ground
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[2] Lee SL, Ramaswamy SD, Aziz MA, Das
Gupta NC, N. C. & Karunaratne GP.
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Post- Vienna Conference on Geotextiles,
Singapore 1987; vol. 2: p. 238-58.
[3] Nguyen TT, Indraratna B. Hydraulic behaviourof parallel fibres under longitudinal flow: anumerical treatment. Canadian Geotechnical
Journal 2016.
[4] Kim JH, Cho SD. Pilot scale field test for
natural fiber drain. Geosynthetics in civil andenvironmental engineering. In: Li G., Chen Y.,
Tang X., Editors. Springer Berlin Heidelberg;
2009. p. 409-414.
[5] Asha BS, Mandal JN. Absorption anddischarge
capacity
tests
onnaturalprefabricated vertical drains. Geosynthetics
International 2012; 19: 263-271.
[6] Lee SL, Karunaratne GP, Ramaswamy SD,
Aziz MA, Das Gupta NC. Natural geosynthetic
drain for soil improvement. Geotextiles andGeomembranes 1994; 13(6–7): 457-474.
[7] Asasutjarit C, Hirunlabh J, Khedari J,
Charoenvai S, Zeghmati B, Shin UC.
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coconut
coir-based
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Building Materials 2007; 21(2): 277-288.
[8] Som N, Chattopadhyay BC, Sanyal T. A
manual on use of jute geotextiles in civil
eingieering, 4th Edition. Jute Manufactures
Development Council, Kolkata, India 2009.
[9] Saha P, Roy D, Manna S, Adhikari B, Sen R,
Roy S. Durability of transesterified jutegeotextiles. Geotextiles and Geomembranes
2012; 35: 69-75.
[10] Dent DL, Pons LJ. A world perspective on acidsulfate soils. Geodarma 1995; 67: 263-276.
[11] Fitzpatrick R, Shand P. Inland acid sulfate soil
systems across australia. In: CRC LEME Open
File Report 249. CRC LEME. Perth. 2008. p.
1-309.
[12] Inglett PW, Reddy KR, Corstanje R. Anaerobic
soils. Encyclopedia of Soils in the Environment.
In: Hillel D., Editor. Elsevier: Oxford; 2005. p.
72-78.

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GEOMATE- Bangkok, Nov. 14-16, 2016[13] Barron RA. The influence of drain wells on the
consolidation of fine-grained soils. TransASCE 1948; 113(2348): 718-54.
[14] Hansbo S. Consolidation of fine-grained soils
by prefabricated drains and lime column
installation. In: Proceedings of 10th
International Conference on Soil Mechanics
and Foundation Engineering, Stockholm, A.A.
Balkema, Rotterdam, The Netherlands 1981;
vol. 3: p. 677-682.
[15] Indraratna B, Rujikiatkamjorn C, Sathananthan
I. Analytical and numerical solutions for asingle vertical drain including the effects ofvacuum preloading. Canadian Geotechnical
Journal 2005; 42(4): 994-1014.
[16] Indraratna B, Sathananthan I, Rujikiatkamjorn
C, Balasubramaniam AS. Analytical andnumerical modeling of soft soil stablization by
prefabricated vertical drains incorporatingvacumm preloading. International Journal of
Geomechanics, ASCE 2005; 5:2(114).

[17] Means JE, Kermit Cromack J, Macmillan PC.
Comparison of decomposition models usingwood density of Douglas. fir logs. Canadian
Journal of Forest Research 1985; 15(6): 10921098.
[18] Gamage NPD, Asaeda T. Decomposition andmineralization of Eichhornia crassipes litter
under aerobic conditions with and withoutbacteria. Hydrobiologia 2005; 541(1): 13-27.
[19] Manzoni S, Piñeiro G, Jackson RB, Jobbágy
EG, Kim JH, Porporato A. Analytical models
of soil and litter decomposition: solutions formass loss and time-dependent decay rates. Soil
Biology and Biochemistry 2012; 50: 66-76.
[20] Indraratna B, Redana IW. Development of thesmear zone around vertical band drains.
Ground Improvement 1998; 2(4): 165-178.
[21] Kim R, Hong SJ, Lee MJ, Lee W. Timedependent well resistance factor of PVD.
Marine Georesources & Geotechnology 2011;29(2): 131-144.

161

Sixth International Conferenc e on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-4-5 C3051

A STUDY ON INJECTING MICRO BUBBLE WATERMIXED WITH CRUSHED SILICA
Yukoh Hasunuma1, Kentaro Uemura1
Takamitsu Sasaki , Koichi Nagao3, Naoaki Suemasa1 and Shunsuke Shimada4
1
Tokyo City University Graduate Division, Japan; 2Kyokado Engineering co.,Ltd, Japan;
3
Sato Kogyo, Japan; 4 Kyokado co.,Ltd, Japan
2

ABSTRACT
Liquefaction has been a serious problem in Japan for a long time and many technical countermeasure methods
have been developed until now. However, it is still difficult to improve the ground in a narrow place such as
residential land. This is because most of the methods generally used for liquefaction countermeasures need bothlarge-scale construction machinery and a high cost per unit area.
Therefore, it has been demanded to develop a method with a low cost and better operability. In this study,
Bubble White was proposed as a new grouting material for a chemical grouting method, which is a mixture ofsilica micro-particles and micro bubble water. This report describes the result of injection experiments where silica
micro-particles were injected into sand specimens in order to verify the permeation characteristics of microparticles. As a result of the experiments, it was confirmed that the permeability of micro-particles in soil changed
depending on both injection speed and amount.
Keywords: Micro-particles, Micro bubble, Liquefaction countermeasures
INTRODUCTION
Japan is one of the countries where earthquakes
occur most frequently in the world. Large earthquakes
in the past caused a lot of damage directly to existing
structures such as shear failure of reinforced concrete
bridge piers. In particular, liquefaction damage has
become a major social issue after the 2011 off the
Pacific coast of Tohoku Earthquake.

this study, therefore, a low-cost material, Bubble
White was focused on as a grouting material for achemical grouting method with excellent operability.
Bubble White is able to be obtained only by mixing
inexpensive silica micro-particles and micro bubble
water.
New grouting material Bubble White
Photo. 2 shows micro bubbles. As thecharacteristics of the micro bubbles, the surface areaof micro bubbles is quite large in comparison withthat of a normal air bubble with the same volume.
Also, the bubbles have an advantage that they tend torise slower and accordingly diffuse horizontally. The
micro bubbles can reduce both saturation degree andpore water pressure in the ground during anearthquake [2].

Photo. 1 Liquefaction damage overview
Photo. 1 [1], taken after the earthquake, shows thedamage situation of underground structures. Groundsubsidence occurred due to liquefaction andaccordingly the manholes buried under the groundcame out from the ground to the surface. In addition,
sand boil phenomena were observed in several areas,
where sand in the ground gushed on the ground. Oneof the reasons why such widespread damage wascaused was that the liquefaction countermeasures forexisting structures have not been pushed forwardsufficiently on account of a high cost performance. In

Photo. 2 Micro bubbles
162

GEOMATE- Bangkok, Nov. 14-16, 2016

CRUSHING EXPERIMENT OF SOLID SILICA

While silica micro-particles are obtained bycrushing solidified silica sol which is made byneutralizing alkali contained in silica glass with acid.
By injection of the silica particles, the density of theground increases. In addition, the silica particles are
viscous enough to stay in the voids of the ground.
They can increase the density of the ground efficiently.
By combining these excellent materials: the microbubble water and the silica micro-particles, it was
expected to increase the effectiveness of preventing
ground from liquefaction [3].

Conditions of Experiment
A problem of silica micro-particles as a grouting
material was that their grain sizes were too large to be
injected into the ground. When a grain size is large,
clogging tends to occur in ground voids at the earlystage of injection and prevent from injecting
uniformly in the ground. In this study, therefore, twodifferent types of solid silica were examined, whichwere cured under different conditions. Their grain
sizes were measured after being crushed in order tostudy grain refining mechanism. Table 1 and Fig. 2
show the test conditions of samples and the grain size
accumulation curves of micro-particles, respectively.

Table 1 Test conditions of samples

casesamplechemical liquid
concentration
curing timeheat curing time
crush time

Photo.3 Solid silica
Photo. 3 shows solid silica. Because the solid
silica was too coarse to be injected into a granularmaterial such as the ground, it was necessary to crush
solid silica into infinitesimal grains by a crushingdevice.

case1
case2
solid silica
12%
7days
120min50min-

Crushing device

solid silica + water

Fig. 1 Crushing device of fine particle

Pump

Fig. 2 Grain size accumulation curves of
microparticles

Fig. 1 shows a crushing device of fine particles.
The crushing device attached to the discharge port ofa pump was used to crush solid silica and generate
micro bubbles. The crushing device has a structure
which produces a turning flow and releases a pressure
when a liquid passes. Silica micro-particles and micro
bubbles were produced by a shear force generated by
the device when a mixture of water with both solid
silica and air released a pressure. Incidentally, airdissolved water was used for the experiments.

Results of Experiment
As a result of comparison between Case1 andCase 2, it was confirmed that Case 1 provided withheat curing was able to be crushed into finer grains
than Case 2. It was assumed that the gel structure wasdisrupted by the effect of volume expansion andcontraction caused by both heat curing andtemperature change after heat curing, and accordingly

163

GEOMATE- Bangkok, Nov. 14-16, 2016solidified silica body became easy to be crushed. On
the contrary, when provided with heat curing until
being dried completely, the solid silica became too
hard to be crushed. It was revealed from the result thatthe crushing efficiency of solid silica increased byproper degree of heat curing. However, since aparticle size of solid silica was still too large to be
used as an injection material, it is necessary tocontinue pursuing optimum curing and crushingmethod.

①injection material②specimen(saturation state)③drainage①③②

Samples for ExperimentFig. 3 and Fig. 4 show the grain size accumulation
curves of silica sands and those of micro-particles,
respectively, which were used for one-dimensional
injection experiments and undrained triaxial tests.

Fig. 5 Schematic of injection experiment
Table 2 Conditions of experiment
case

case1

case2

silica sand#5

#6
microparticle①

diameter

5.0cm
50cm

concentration of
injection material
injecting pressure

case4
#7

injection material

heightFig. 3 Grain size accumulation curves of Silica Sand

case3

30cm

4.0%
20kPa

relative density

20kPa

40kPa

40kPa

60%

ONE-DIMENSIONAL INJECTION EXPERIMENT
Conditions of Experiment
The injection experiment was conducted in order
to confirm the permeability of micro-particles which
may differ depending on sand types. A long specimen
was prepared for the experiment by combining
separable columns with the height of 10.0 cm. Fig. 5
and Table 2 show a schematic and the conditions ofinjection experiment, respectively.
The sand specimen with 5.0cm in diameter and50.0cm in height was prepared by the air-pluviation
method at a relative density of 60% and completely
saturated. And also, the drainage from the outlet ofthe container was sampled and its material
concentration was measured after drying.

Fig. 4 grain size accumulation curves of micro-particles

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GEOMATE- Bangkok, Nov. 14-16, 2016

Results of ExperimentFig. 7 shows the change of replacement rate withtime. Case 1 and Case 2 showed almost the same
behavior, as they were injected at the same constant
speed. After the replacement rate exceeded 1.0, theinjection volume of Case 4 gradually decreased. Thisis considered that clogging occurred even in the caseof the short permeation distance because the particle
diameter of silica sand #7 was small. In Case 3, theinjection speed was extremely lower than those of theother cases because the flow resistance of the mixture
was larger than those of the other cases when the
micro-particles passed through narrower void ofsilica sand #7. From the results, it seems that thefactors on whether the clogging of the micro-particles
occurs are attributed to the size of sand where themicro-particles pass through, the replacement rate of
the mixture and the distance of the micro-particles
passing through.

Results of experiment are shown in Fig. 6 to Fig.
8. In these figures, the index C / C0 is the ratio of theconcentration of material in drained C to the initialconcentration of material in injected mixture C0. The
term “replacement rate” is defined as the ratio of
injected mixture volume divided by total void volume
in the specimen.

Fig. 6 Relationship between replacement rate andconcentration ratio
Fig. 6 shows the relationship between
replacement rate and concentration ratio. The
concentration ratio increased in the replacement rateof 1.0 in all cases except Case 3. On the other hand,the concentration ratio of Case 3 increased after the
replacement rate exceeds 1.0. This result is thought tobe caused by using silica sand C for Case 3, which isfiner in particle size than sands for Case 1 and Case 2.
Silica sand #7 is likely to clog micro-particles
because the void diameter is smaller than the othersands. Therefore, it is considered that the increase inconcentration ratio was suppressed because theamount of micro-particles discharged was reduced. In
Case 4 with silica sand #7, however, its upward trendof concentrations was similar to the other cases.
Because the permeation distance was shorter thanthose of the two cases, it was found that the effect ofparticle size of sand was balanced by the influence of
permeation distance.

Fig. 8 Relationship between concentration ratio
and permeation distanceFig. 8 shows the distribution of concentration
ratio of silica particle in pore water with permeation
distance. As the concentration ratios in all cases
except Case 3 are plotted around 1.0, which means
injected materials were uniformly injected anddeposited. On the other hand, the concentration ratio
in Case 3 has a large variation. It is presumed that the
injected material is not uniformly injected in the casethat injection speed is slow. In other words, there is apossibility that the concentration ratio became locallyhigh in the position where clogging occurred. Photo.
4 shows clogging of the specimen.

Fig. 7 Relationship between injecting time andreplacement rate

Photo. 4 Clogging of the specimen
165

GEOMATE- Bangkok, Nov. 14-16, 2016

UNDRAINED CYCLIC TRIAXIAL TEST

Table 4 Conditions of experiments

Conditions of Experimentcase

A series of cyclic triaxial tests were performed to
study the liquefaction strength on the specimen
improved by injecting micro-particles and micro
bubbles. Table 3 and 4 show the conditions ofexperiments. All of the specimens except Case 2 wereprepared at the relative density of 60%. Case 1 andCase 2 were conducted to obtain the liquefaction
strength of unimproved sand specimens at the relative
density of 60% and 70% for comparison withimproved sand specimens. As for Case 3 to 6,
grouting material was injected after the specimens
were fully saturated. Incidentally, silica microparticles used in Case 3 and 4 had an average diameterof about 20μm. In addition, a localized necking wasobserved in the specimen at the situation of DA= 5%
during the test of Case 6 where the bubble white wasinjected. Photo. 5 shows localized necking of thespecimen. The necking failure was thought to be
shear failure at the tensile side in the triaxial test
rather than liquefaction. Therefore, the test results
were estimated at the double amplitude of axial strain:
DA=3% not to cause the localized necking failure in
all cases.

case1
case2
case3
case4

injection material-

microparticle②

case5

microparticle①

case6

microparticle①+
Micro Bubble

Concetration of Injection
injection material pressure
6%
20kPa
12%

Fig. 9 Liquefaction strength curves
Results of ExperimentFig. 9 shows the liquefaction strength curves ofboth improved and unimproved sand specimens.
The relative density of each improved specimen
was calculated by a difference between specimen
weights before and after the injection experiment
which was carried out in another process. The relative
density of Case 3 was 63% and those of Case 4 and 5
increased to 69%. As for Case 3, however, theliquefaction strength was almost the same as those of
unimproved sands. Although the liquefaction strengthof Case 4 and 5 increased, it was lower than that ofunimproved sand whose relative density was 70%. It
was expected that the soil skeleton was notstrengthened by the injected particles because theinjected particles didn’t adhere to the contact portionof the soil skeleton. From this experiment, it was
confirmed that a portion of the injected particles
strengthened soil skeleton. Furthermore, the effect ofsuppressing liquefaction increased more by injecting
Bubble White than injecting only fine particles.

Photo. 5 Localized necking of the specimen
Table 3 Conditions of specimens
samplesilica sand #5

diameter

5.0cm

height10.0cmrelative density

60%,70%(case2)

166

GEOMATE- Bangkok, Nov. 14-16, 2016

CONCLUSIONREFERENCE

Solid Silica was found to be crushed efficiently
by moderate heat curing.
The permeability of micro-particles was affectedby not only penetration speed and injection volume
but also the particle size of sand injected. Especiallywhen the particle size of the sand was small,permeability was greatly affected by permeation
distance.
Liquefaction strength was found to increase by
the injection of micro-particles and Bubble White.
However, the liquefaction strength of improved sand,whose relative density increased by injection, became
lower than that of unimproved sand with same
relative density. Therefore, it was cleared that evenrelative density of specimen increased by injecting
particles, not all the injected particles increasedliquefaction strength. Moreover, since liquefaction
strength increased more by injecting Bubble White
than injecting only micro-particle, it was figured outthat Bubble White was effective to suppressliquefaction.

[1]Katori City website, http://www.city.katori.lg.jp
2016/01/07[2]Koichi N, Naoaki S, Tosiyuki K, Saeka
Y,‘‘Applicability of soil improvement with using
Micro-Bubbles mixing water against soilliquefaction’’

J.JSCE

Ser.C

2015

Vol.71 No.4 pp.395-406. (in Japanese)[3]Shota M, Kentaro U, Naoaki S, Takamitu S,
Koichi N, ‘‘Ultrafine particles of silica grouting
method for liquefaction countermeasure ’’The
Japanese Geotechnical Society, 2015, p.307. (inJapanese)167

Sixth International Conferenc e on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-4-5 C3051

SOIL BEARING CAPACITY REFERENCE FOR METRO MANILA,
PHILIPPINES
Jonathan R. Dungca1, Ismael Concepcion Jr.1, Moises Christian Mickail Limyuen1,
Terence Oliver See1 and Marion Ryan Vicencio1
1

Gokongwei College of Engineering, De La Salle University, Philippines

ABSTRACTThis study focuses on the analysis of the soil bearing capacities of the various cities and municipalities ofMetro Manila, Philippines. The allowable soil bearing capacities to be used for foundation design were
calculated through various theories and studies using geotechnical parameters, such as relative density and angle
of internal friction. Standard Penetration Test (SPT) results were used to estimate these geotechnical parametersin order to obtain a good approximation of the soil‘s bearing capacity.
Because of economic constraints, not all low-rise construction projects choose to perform soil exploration.
Due to this, soil data are usually lacking and may cause problems when designing shallow foundations of thesekinds of structures. In line with this kind of situation, the study can help engineers in designing shallow
foundations by providing them a reference of the allowable bearing capacity of any area within Metro Manila.
This will be able to give them a good idea of the soil‘s strength in supporting shallow foundations.
The allowable bearing capacity of the soil shown in the reference is obtained from collected borehole data
within Metro Manila and by using several geotechnical engineering theories. Contour maps of the bearing
capacities are then made in order to provide an overview of the soil‘s bearing capacity for shallow foundations.
A Geographic Information System (GIS) software database was also made so as to store all the borehole
location‘s data as well as serving another basis for estimation. This can be updated whenever new data isavailable.
Keywords: Soil Bearing Capacity, Foundation Design, Standard Penetration Test, Geotechnical Properties ofMetro Manila

provide structural engineers the estimated allowablesoil bearing capacities at any point in Metro Manila
Philippines through the use of contour maps.

INTRODUCTION
Foundation design requires engineers tounderstand how the soil interacts with thefoundations. But, foundations are situatedunderground, wherein engineers cannot explicitly
describe the interactions of the soil undergroundwithout conducting some tests. As soil exploration isa very costly test to conduct, engineers cannot
always perform these tests; therefore, they rely on
previous explorations done by their peers that areclose to the project site to approximate the value forthe bearing capacity for the aforementioned.
Since soil exploration is too costly and mere
guessing will not suffice when making foundations,the author decided to come up with a way to fill inthe dire needs of engineers. Using the borehole logs
available within the Metro Manila, Philippines, thestudy analyzed the SPT borehole logs, calculated therequired soil parameters and compiled everything
into a reference that shows the bearing capacities of
the covered areas.
The study aims to create a reference that will

METHODOLOGYWith Metro Manila having an approximate sizeof 624.33 square kilometers, a density of one
borehole log per square kilometer was used todescribe the geotechnical characteristics andpossible foundation design parameters of the saidarea. Borehole logs were collected for a total of 486
locations all over Metro Manila.
The amount of borehole logs alone is not the
only criterion in gathering data for the study. It isjust as important as, that the locations of theborehole logs be properly distributed. To check theirdistribution, each of the locations of the boreholelogs was plotted in a map of Metro Manila. After
this, the distribution was visually inspected and theareas that needed more data were determined. Aside
168

GEOMATE- Bangkok, Nov. 14-16, 2016

allowable bearing capacities at different locations in
Metro Manila. With the ever growing technology indata processing, it has become more convenient toshow the reference through the use of a Geographic
Information System (GIS) program. The map ofMetro Manila was first loaded into the GIS program.
After this, the locations and all the relevant, datasuch as the bearing capacities, were placed insidea .csv file.

from these, borehole logs that seemed erroneous
were removed and disregarded.
In properly designing shallow foundations, thegeotechnical characteristics and allowable bearing
capacity of the soil must be known. This is because
the design would largely depend on the strength andthe behavior of the soil. The bearing capacities arecomputed using the SPT N values found in theborehole logs which were corrected using theprocedures discussed in [1]-[4], also shown inEquation 1.
𝑁𝑁60 =

𝐸𝐸𝑚𝑚 𝐶𝐶𝐵𝐵 𝐶𝐶𝑆𝑆 𝐶𝐶𝑅𝑅 𝐶𝐶𝑁𝑁 𝑁𝑁
0.6

ANALYSIS AND DISCUSSION
Metro Manila or the National Capital Region of
the Philippines is composed of 16 cities and 1
municipality. It is bounded by the province of
Bulacan in the North, Manila Bay at the West, Rizal
Province and Laguna Bay in the East, and Cavite
Province in the South. It has a total land area of
597.47 km2.

(Eq. 1)

Where: N60 is the corrected SPT-N value (blows/ft),
Em is the hammer efficiency, CB is the borehole
diameter correction, CS is the sampler correction, CR
is the rod length correction, CN is the overburdenpressure correction and N is the SPT-N recorded in
the field.
The corrected SPT, N60, values were then used tocompute for various geotechnical parameters such as
relative density, undrained shear strength and angle
of internal friction using different correlation factors[5]-[9]. As such, the group computed the ultimate
bearing capacity. The Terzaghi‘s and Vesic‘sbearing capacity formulas, shown on Equations 2
and 3 respectively, were used to achieve this [10][12].

Soil Characteristic
Manila used to be a submerged area at one timein the geologic past. Intermittent volcanic activities
followed and after which, volcanic materials were
deposited. During the intervening period ofinactivity, transported sediments were deposited on
top of previously-laid volcanic materials. Thus,
alternating beds and transported sediments became acharacteristic feature of the geologic deposit. [13][14]Available geologic information about Metro
Manila and the areas surrounding it indicated that
Quaternary volcanic rocks generally known as the
Guadalupe Formation, locally known as “adobe”, isthe predominant rock unit underlying it. It consistsof the Lower Alat Conglomerate Member and theUpper Diliman Tuff Member. The Diliman Tuff
includes the tuff sequence in the Angat-Novaliches
region and along Pasig River in the vicinity of
Guadalupe, Makati and extending to some areas ofManila and most of Quezon City. Its upper surface
ascends gently from Manila Bay outward Caloocan,
Makati, Mandaluyong and Quezon City in whichrock exposures can be found. The entire sequence isalmost flat lying, thin to thickly bedded and consistof medium to coarse grained vitric tuffs and welded
volcanic breccias with subordinate amounts of
tuffaceous medium to coarse-grained sandstone,
silty and clayey tuffs and tuffaceous conglomerates.
These types of tuff are distinguished from their
textual characteristics. Silty and clayey tuffs are veryfine-grained while conglomerates contain coarse

𝑞𝑞𝑢𝑢𝑢𝑢𝑡𝑡 = 1.2𝑐𝑐𝑁𝑁𝑐𝑐 + 𝛾𝛾𝛾𝛾𝛾𝛾𝑁𝑁𝑞𝑞 + 0.4𝛾𝛾𝛾𝛾𝑁𝑁𝛾𝛾 (Eq. 2)

Where: qult is the ultimate bearing capacity, γ is theeffective unit weight, B is the width of foundation,
Df is the depth of foundation below ground surface,Nc, Nγ and Nq are the Terzaghi’s factors.
′
𝑞𝑞𝑢𝑢𝑢𝑢𝑢𝑢 = 𝑐𝑐 ′𝑁𝑁𝑐𝑐𝑠𝑠𝑐𝑐 𝑑𝑑𝑐𝑐𝑖𝑖𝑐𝑐 𝑏𝑏𝑐𝑐𝑔𝑔𝑐𝑐 + 𝜎𝜎𝑧𝑧𝑧𝑧
𝑁𝑁𝑞𝑞 𝑠𝑠𝑞𝑞 𝑑𝑑𝑞𝑞 𝑖𝑖𝑞𝑞 𝑏𝑏𝑞𝑞 𝑔𝑔𝑞𝑞
+ 0.5𝛾𝛾′𝑁𝑁𝛾𝛾 𝑠𝑠𝛾𝛾 𝑑𝑑𝛾𝛾 𝑖𝑖𝛾𝛾 𝑏𝑏𝛾𝛾 𝑔𝑔𝛾𝛾

(Eq. 3)

Where: sc, sq, sγ are shape factors; dc, dq, dγ are depth
factors; ic, iq, iγ are load inclination factors; bc, bq, bγ
are base inclination factors; and gc, gq, gγ are ground
inclination factors.
A factor of safety of 3.0 was divided to ultimate
bearing capacity to determine the allowable soil
bearing capacity of the soil. The allowable soil
bearing capacities at different locations were thenplotted on to the maps at 1-meter, 2-meters, 3meters , 4-meters and 5-meter depths. Contour maps
were created to visually classify and analyze the169GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Number of Borehole Locations and its
Density

particles. In some areas, this rock formation isoverlain by minimal alluvial deposits which tend to
thicken towards Manila Bay. The high elevation
areas are generally composed of dense sands andtuffaceous clay, while the low-lying areas are
generally composed of loose sands and soft clays[13]- [14].
Topography and Elevation
Figure 1 shows that majority of the Metro Manila
area rest on high elevations. The low lying areas are
the coastal areas.

City or
Municipality

Area
(km2)

ManilaMandaluyong
Marikina
Pasig
Quezon CitySan JuanCaloocanMalabonNavotasValenzuela
Las Pinas
Makati
Muntinlupa
Paranaque
Pasay
Pateros
Taguig
TOTAL

38.55
11.25
21.50
31.00
134.26
5.94
53.33
15.71
10.69
44.59
41.54
27.36
46.70
46.57
18.50
2.10
47.88
597.47

Borehole
Locations
(BH)
39
14
26
24
130
3
15
12
11
9
38
28
17
40
24
8
48
486

Density,
BH/Area
1.00
1.00
1.00
0.77
0.97
0.51
0.28
0.76
1.00
0.20
0.91
1.00
0.36
0.86
1.00
1.00
1.00
81.34%SOIL BEARING CAPACITY

Fig. 1Figure 3 shows the bearing capacity of the region
at 1-meter depth. It can be seen that many areas
where elevation is high have larger bearing
capacities compared to the lower elevated areas. It isobserved that the fault line is directly below the part
where the bearing capacity is high. SPT N values ofthe outer 1-meter depth have a usual value in therange of 2 to 10 for both sands and clays. A reason
why the outer areas have low bearing capacity is due
to sediment deposits that are left by the rivers andcreeks over time.

Surface Elevation Contour

Borehole Locations
Figure 2 shows the location of the data points
that were obtained in the area while Table 1 shows
the number of borehole location in each area of thestudy and density of data per one square kilometer.

Fig. 3Fig. 2

Bearing capacity at a Depth of 1 meter inMetro Manila

The allowable bearing capacities of a depth of 2
meters below the ground surface are shown in

Map of the Data Points in Metro Manila

170

GEOMATE- Bangkok, Nov. 14-16, 2016

Figure 4. It can be seen in the central part of theregion that refusal (SPT N-values greater than 50)
has been achieved. Bearing capacities of the outerportions has also increased but not as muchcompared to the center of the map. The blue areas
represent the places where the soil has hardened to apoint that SPT is not advisable anymore. These blue
areas are very near, if not above, the fault line andthus have a large bearing capacity.

Fig. 6

Fig. 4

Bearing capacity at a Depth of 4 meters in
Metro Manila

Figure 7 shows the bearing capacity at a depth of
5 meters below the ground surface. Almost all areas
have reached an allowable bearing capacity of 300
kPa and above, as expected, due to consolidation ofsoil over time.

Bearing capacity at a Depth of 2 meters in
Metro Manila

Figure 5 shows the bearing capacities of a depthof 3 meters below the ground surface. More areas
have now achieved refusal. SPT N values of theareas around the blue areas have increased but theouter areas of Metro Manila still have a small
bearing capacity as compared to the center.

Fig. 7

Fig. 5

Bearing capacity at a Depth of 5 meters in
Metro Manila

Figure 8 shows the geological map of Metro
Manila [13]. Looking at this map, deposits are seenin the western and eastern side of the region, while
the tuff formation is seen in the central region.
With this information, the region can be further
divided into 3 parts: the west coast, the east coastand the central area as shown in Figure 9. The west
coast is composed of cities that are near the Manila
Bay, which are Navotas, Malabon, South Caloocan,
Manila, Pasay, Paranaque and Las Pinas. The east
coast is composed of cities that are near the Laguna
Bay, which are Marikina, Pasig, Pateros, Taguig andBearing capacity at a Depth of 3 meters in
Metro Manila

Figure 6 shows the bearing capacity at a depth of
4 meters below the ground surface. Almost all areas
have a bearing capacity of 300 kPa or greater except
in the northwestern and eastern portion.
171

GEOMATE- Bangkok, Nov. 14-16, 2016

Muntinlupa. The remaining cities of North Caloocan,
Valenzuela, Quezon City, San Juan, Mandaluyong,and Makati are the cities that comprise the centralarea.

Fig. 8ground surface in the inner regions.

Fig. 10 Bearing Capacity Contour Map of West andEast Coast of Metro Manila (depth of 1m)

Geologic Map of Metro Manila [13]Since the surface geology of this area isreported as tuff, it can be generalized that the land ismostly made up of rocks. Comparing rocks to soil,
rocks are generally stronger than soil. In accordance
to this, the bearing capacities of the rocks shouldhave high values compared to soils which is clearly
noticeable in Figure 11. Calculations have shownthat the bearing capacities of the area have a rangeof 200 to 300 kPa.

Fig. 9Division of Cities

Looking at the west coast, as shown in Figure 10,
the surface geology of this area is mostly composedof quarterly alluvial soil. SPT N- values on theseareas are also very low (2-10) at the surface andaverage (20-30) as the soil goes deeper. The east
coast is generally the same compared to the west
coast, having quarterly alluvium soil in the surface
as well as having the same SPT value patterns. SPT
refusal was not achieved below a depth of 5 meters.
Due to the fact that the surface geology of the soil ofthe aforementioned areas is generally loose, theresulting bearing capacities also reflected thecomposition of the soil. Allowable bearing
capacities of the area at shallow depths of 1 to 2
meters have values ranging from 0 to 200 kPa.
The central area of Metro Manila has a different
surface geology compared to its neighboring sides.
Tuff was primarily observed in this area, and rockformations are common below the surface. SPT
blows depend on the amount of deposits that arepresent and increase as the soil goes deeper. SPT
refusal was achieved at shallow depths of 2 to 3
meters in the outer regions and as shallow as theFig. 11 Bearing Capacity Contour Map of Central
Area of Metro Manila at a depth of 1 meter
CONCLUSION
Borehole logs were collected for a total of 486
locations all over Metro Manila, Philippines. Maps
were also created in order to show the locations of
the collected boreholes throughout Metro Manila.
Using an excel program, different geotechnicalproperties of the collected boreholes were computed.
Some examples of this are the unit weight and theangle of internal friction. Also using the same excel
program, bearing capacity of the 16 cities and 1
172

GEOMATE- Bangkok, Nov. 14-16, 2016

municipality were estimated and evaluated per meter
depth until a depth of 5 meters. Contour maps of theallowable bearing capacities at each of the depths
were also made and analyzed. Based on the resultsof the study, it was found out that several locationsin Metro Manila were composed of soil withallowable bearing capacities suitable for shallow
foundations.
Cities which are near the bodies of water, such
as Manila, Navotas, and Marikina, have low bearing
capacities. Thus, the use of shallow foundations onthese areas is recommended only for structures thathave lower design loads, such as residential houses.
As the soil is not capable of carrying heavy loads,
using shallow foundations for high rise buildings
and other large structures should be avoided or adeep foundation is recommended. According toresearch [13], the western part of Metro Manila wasonce underwater. Due to this, some of the soils in
the area contained seashells. Also, the surface
geology of the western and eastern area is composed
mostly of quarterly alluvium, a loose type of soilthat is in its early ages. Also, it should be noted thatthe river system determines the direction of thesediments being washed out from its origins.
Marikina River flows from north to south. Pasig
River then carries the river water from Marikina
River, splitting it into two: one in the west direction
with the Manila Bay as its last stop, while the otherin the east direction with the Laguna Bay as its
endpoint. Because of this, sediments from highelevation areas were deposited by rivers and creeks
into low lying areas along their paths. This happens
because water slowly scours the soil and carries it to
the areas where the slope is mild. Gravity then takes
effect and sediments are left behind, resulting in acomposition that is generally loose and soft forsands and clays.
Cities with rock formations beneath the surface,such as Quezon City, North Caloocan, andMuntinlupa, have soils with high bearing capacities
at shallow depths. It is recommended to place thefoundations on these refusal levels since it is morethan capable of carrying loads that are suited forshallow foundations. Nevertheless, caution must be
taken when placing structures in these areas, as theValley Fault System is nearby, making the area
prone to earthquakes.

teJohn Wiley, 1943. Terzaghi K, “TheoreticalSoil Mechanics”, John Wiley, 1943.
[2] Clayton
CRI
“
SPT
EnergyTransmission:Theory,
Measurement,
andSignificance”, Ground Engineering, Vol. 23, No.
10, 1990, pp 35-43.
[3] Skempton AW “ Standard penetration Test
Procedures and the Effects in Sands ofOverburden Pressure, Relative Density, Particile
Size, Aging and Over Consolidation”
Geotechnique, Vol 36, No. 3 1986, pp. 425-447.
[4] Tokimatsu k and Seed HB “Simplified
Procedures for the Evaluation of the Settlements
in Clean Sand. Report No. UCB/EERC-84/16.
University of California , Berkely, CA, 1984.
[5] Das B. “Fundamentals of GeotechnicalEngineering, 3rd Ed., Toronto, Canada,Thomson Learning, 2008
[6] Coduto DP “ Foundation Design Principles andPractices, 2nd Ed., New Jersey, USA, Rentice
Hall, 2001.
[7] Sowers G. and Sowers G. “Introductory soil
Mechanics and Foundation, New York, USA
Macmillan, 1961.
[8] Gibbs HJ and Holtz WG “Research on
Determining the Density of Sands by Spoon
Penetration Testing, In the Proceedings of the
4th International Conference on Soil Mechanics,
Vol. 1, London, 1957, pp. 35-39.
[9] Kulhawy FH and Mayne PW “Manual on
Estimating Soil Parameters for Foundation
Design. Report No. EL-6800, Electric Power
Research Institute, Palo, Alto, CA, 1990.
[10] Terzaghi K, “Theoretical Soil Mechanics”, John
Wiley, 1943. Terzaghi K, “Theoretical Soil
Mechanics”, John Wiley, 1943.
[11] Vesic’ AS “Analysis of Ultimate Loads of
Shallow Foundations,” ASCE Journal of theSoil Mechanics and Foundation Division, 1973,
Vol. 99, No. SM1, pp. 45-73.
[12] Vesic’ AS “Bearing Capacity of Shallow
Foundations,” Foundation Engineering
Handbook. 1st ed., p 121-147, Winterkorn, Hans
F. and Fang, Hsai-Yang, eds., Van Nostrand
Reinhold, New York., 1975.
[13] Oca G. “The Geology of Greater Manila and Its
Bearing to the Catasthropic Earthquake ofAugust 2, 1968” the Philippine Geologist, Vol
22 No. 4, pp. 171-192.
[14] Tan R. “Engineering Properties of Manila
Subsoil” Quezon City Philippines: University of
the Philippines Press, 1985.
[15] Bureau of Mines and Geo-Science,” Geology
and mineral resources of the Philippines”, 1981.

REFERENCES[1] Dungca J R and Chua RAD, “Development of a
Probabilistic Liquefaction Potential Map for
Metro Manila”, Int. Journal of GEOMATE,
April 2016, Vol. 10, No. 2, pp. 1804-1809

173

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

VALIDATION OF METHOD FOR ESTIMATING LONG-TIME STRENGTH OFCHEMICAL GROUT COMPOSITION WITH HEAT CURING
Shunsuke Takiura1, Naoaki Suemasa1 and Takamitsu Sasaki2
Urban and Civil Engineering, Tokyo City University, Tokyo, Japan
2
Research and Development, Kyokado Engineering Co., Ltd, Tokyo, Japan
1

ABSTRACTThe Great East Japan Earthquake on March 11, 2011 caused terrible liquefaction damage especially inreclaimed lands and soft grounds. Under the circumstances, there has been a growing demand for a liquefaction
countermeasure in Japan. In this study, a chemical grouting method was focused on as a liquefaction
countermeasure method because it is excellent in performance compared to other methods. Although the chemical
grouting method used to be only a temporary and supplementary construction method, it has become one of themost effective liquefaction countermeasures in recent years since the performance of the chemical solution has
been improved and is capable to make long-term performance in the ground. However, the mechanism of thechemical solution has not been clarified yet, because it would require quite a long time to verify the long-term
durability of the improved ground. It is urgent and essential to find out an efficient method to verify it in a shortperiod of time. In this study, the possibility to predict the change of long-term strength was examined by a shortterm laboratory test where heat curing was provided the improved ground model in a thermostatic tank. Here, theacceleration mechanism by heat curing is based on the chemical reaction rate theory of Arrhenius. The predictedvalues calculated from the Arrhenius equation showed similar strength change to the measured values obtained
from the test. As a result, it was confirmed that it was possible to calculate the future change of the durability withthis method.
Keywords: Chemical grouting method, Long-term durability, Heat curing
the sand against earthquake vibration [2]. The
advantages of the method are less vibration and lessnoise than other methods during the construction. In
addition, this method provides small equipment forspace saving, which can drill a hole not only in a
vertical but also in an oblique and a curved lines.

INTRODUCTION
Chemical Grouting Method for
Liquefaction Countermeasure
The Great East Japan Earthquake disaster
happened on March 11, 2011 and the maximum
acceleration of 2933 gal was recorded in Kurihara
city in Miyagi prefecture. Liquefaction occurred inmany regions from Tohoku to Kanto district and thetotal damage throughout Japan reached42 square
kilometers. The liquefaction countermeasures using achemical grouting method were performed in several
places along the Pacific coast in 2008 after Niigata
Chuetsu-oki earthquake in July 2007. It was revealed
through the damage survey of the 2011 Earthquakethat the chemical grouting method was quite effective
as a liquefaction countermeasure [1].

Durability of chemical improved body
A human-initiated accident took place in 1974
where the groundwater was contaminated at the
ground improvement work injecting the materials of
Acrylic Alumite origin. The former Ministry ofConstruction prohibited the useof themacromolecule-based infusion materials promptly
after the accident and only water glass-based grout
materials have been used since then. However, it iswell known that the durability of water glass-based
grout infusion materials is not strong enough to
support the ground for a long time that they are
generally used only for temporary constructions [3].

Improvement Principle and Characteristic
The improvement principle of the chemical
grouting method is to replace pore water in sandskeletons with a gelatinous silica compound in orderto prevent both contraction of the sand skeletons anda rise in excess pore water pressure caused by theseismic vibration. In addition, the silica compound
gives cohesion between sand particles andaccordingly increases the cyclic shear resistance of

Recently, the effect of water glass-based grout
infusion materials has been improved through some
new technology, and consequently a new type ofdurable grout material has been developed. However,there are still problems for this new material. One of174GEOMATE- Bangkok, Nov. 14-16, 2016

these is that most of previous studies on the chemical
grouting have focused on the development or the
choice of durable infusion materials. Furthermore,
sufficient proof experiments for the material have notbeen carried out because it has been only ten andseveral years since the permanent grout material wasdeveloped. It is, therefore, essential to establish apractical way to estimate the strength change of the
improved soil with elapsed time [3].

Table 1.

Physical properties of Toyoura sandDensity test ofsoil particle(g/cm3 )Maximumvoid ratioemax

Minimumvoid ratio
emin

2.634

0.986

0.621

100

Passage mass percentage %Purposes of this study
Athough a chemical grouting method of
construction has been proved to be effective forliquefaction countermeasures, the long-term strengthof the improved soil has been unexplained. Therefore,
the method in which the long-term strength is able tobe verified in a short term is suggested in this study.
In the study, a heat and cure process have been
provided for the chemical improved specimens in athermostatic tank in order to promote the chemical
reaction. The strength change of the specimens
applied the process of heating and curing for 1000days is reported in this paper.

80
60
40

20
0
0.001

Fig.1

0.01

0.1
Sieve (mm)1

10Grain size accumulation curve of
Toyoura sand

EXPERIMENTAL PROCEDURE

Active colloidal silica
Volime change %Physical characteristics of sample
Toyoura sand was used for this experiment. Table
1 shows the maximum and the minimum densities of
Toyoura sand and Fig. 1 shows the result of grain size
analysis.

Organic water glass

Silica sol

Characteristics of grout material
The grout material used for the experiments wereactive colloidal silica with 29% of concentration inweight and silica sol with 6%. The active colloidal
silica is based on the active silica, which is obtainedby removing alkali from water glass by an ion
exchange technique in order to exclude its
deterioration factor. Therefore, the active colloidal
silica is considered to become a sort of permanent
grouting materials that could keep its effectivenesswithout eluviation of silica for a long time. While, thesilica sol is based on the non-alkaline water glass
solution which is obtained by neutralizing alkali of
the water glass with acid [4]. As shown in the Fig 2,
the active colloidal silica has hardly changed involume for a long time. On the other hand, the silica
sol has remarkably changed (decreased) in volume
immediately after it was solidified. And also, theunconfined compression strength of the improved soil
tended to gradually decrease. It seemed to be because
the cohesion of the gel decreased or disappeared by
the decrease in volume.

Time (day)Fig.2 Volume change of plain gel(Geo-grouting development Organization: HP)

Methods of making a specimen and curing
Two kinds of specimens, sand gel and plain gel,were prepared as follows. For the sand gel, Toyoura
sand was air-pluviated in a mold of 5 cm in inside
diameter so as to be the sand specimen with 60% of
the relative density. The sample was set 10cm inheight by putting a weight on top of the sample andtaping a side of the mold with a wooden hammer.
Then the chemical grout was injected into a specimen
deaerated in a vacuum pump. After confirming thatthe specimen was fully solidified after 24 hours, thespecimen was removed from the mold. Photo. 1 andPhoto. 2 describes the situation of injecting the175

GEOMATE- Bangkok, Nov. 14-16, 2016

chemical grout solution. For the plain gel, 200 ml ofchemical grout was poured into a glass container with
5cm in inside diameter and the specimen molded by
the solidification of the grout was removed from thecontainer.

Table 2.

Experiment case

Grouting

Test piecesand gel

Activecolloidal silica

A process of curing was carried out under theconditions of three different temperatures of 25, 45
and 50 degrees Celsius. All of the specimens wrappedin a wrap film were covered with wet papers and put
in an airtight container to prevent from being dried.
As for the curing at 25 degrees Celsius of temperature,
the specimen was cured in the room where thetemperature was constantly controlled at 25 degrees
Celsius with an air-conditioner. For the heat curing,
the specimens in the containers were set in theconstant temperature tanks whose temperatures were
continuously adjusted at 45 and 55 degrees,
respectively. An experiment case is shown in Table 2.

plain gelsand gel

Silica sol
plain gelCuringtemperature

site

25℃
45℃
25℃
45℃
25℃
45℃
55℃
25℃
45℃
55℃

A
B
C
D
E
F
G
H
I
J

RESULTS AND DISCUSSION
Unconfined compression test result

Photo.1

Fig. 3 shows the change of unconfinedcompression strength for the sand gel specimen of theactive colloidal silica with time and temperature. The
specimen of heating curing showed higher
unconfined compression strength than that ofordinary temperature curing. Fig. 4 shows the changeof unconfined compression strength for the plain gel
specimen of the active colloidal silica. In the same
way as the sand gel, the specimen of heating curing
showed higher unconfined compression strength thanthat of ordinary temperature curing. From the result
mentioned above, a reaction acceleration effect by theheating curing was confirmed.

Infusion situation

Unconfined compression strength (kN/m2)

1200
A
1000

B

800
600
400
200

0
0

Fig.3

Photo.2

Infusion situation

176

200

400

600
Time (day)

800

1000

Unconfined compression strength ofactive colloidal silica sand gel

1200

GEOMATE- Bangkok, Nov. 14-16, 2016

160

140

C

120

D

Unconfined compression strength (kN/m2)

Unconfined compression strength (kN/m2)

160

100
80
60
40
20
0

0

200

400

600

800

1000

1200

I
120

80
60
40
20
0
0

Unconfined compression strength ofactive colloidal silica plain gel

Fig.6

Fig. 5 shows the change of unconfinedcompression strength for the sand gel of the silica sol.
The specimen of heating curing showed lower
unconfined compression strength than that ofordinary temperature curing. In addition, the sand gel
specimen of the silica sol had gradually decreased instrength. Similar results were obtained in all cases forthe silica sol. Fig. 6 shows the change of unconfinedcompression strength for the plain gel specimen of thesilica sol. In the same way as the sand gel, thespecimen of heating curing showed lower unconfinedcompression strength than that of ordinary
temperature curing. The specimen of ordinary
temperature curing had gradually increased instrength. On the other hand, as for the specimen ofheating curing, the strength had gradually decreased
after having increased at an initial stage.

400
Time (day)

600

800

Unconfined compression strength ofsilica sol plain gel

Fig. 7 and 8 show the changes in the rate ofvolume change of specimen for the plain gel of theactive colloidal silica and silica sol over the time,
respectively. As for the active colloidal silica, the rateof volume change was constantly near -2.5% in bothcases of ordinary temperature curing and heating
curing. While, as for the silica sol, the rate of volume
change in the case of heating curing grew larger thanthat in the case of ordinary temperature curing. In
addition, when comparing the specimens of heating
curing, the amount of volume change was larger at ahigher temperature. As shown in Fig. 2, the silica sol
tends to change in volume easily. Therefore, itappeared that heating curing promoted the volume
change more effectively than ordinary temperature
curing.

E
1000

200

Volume change of plain gel

1200
F
G

800

0

200

400

600

800

1000

1200

0.0

600

-5.0

400

Volume change (%)

Unconfined compression strength (kN/m2)

J

100

Time (day)

Fig.4

H

140

200
0
0

200

400

600

800

1000

D

-15.0
-20.0
-25.0
-30.0

Time (day)

Fig.5

C

-10.0

-35.0

Unconfined compression strength ofsilica sol sand gel-40.0Time (day)

Fig.7

177

Volume change in active colloidal silica
plain gel

GEOMATE- Bangkok, Nov. 14-16, 2016

200

400

600

5

800

Change rate of the strength (⊿kPa/⊿t)

0
0.0

H

-5.0
Volume change (%)

I
-10.0

J

-15.0
-20.0
-25.0
-30.0
-35.0

Silica sol

4
3.5
3
2.5
2
1.5
1
0.5
0
3.00

-40.0
Time (day）Fig.8

Active colloidal silica

4.5

55℃

3.10

3.20

45℃

3.30

3.40

25℃

1000/T

Volume change in silica sol plain gel

Fig.9

Arrhenius plot of sand gel

Unconfined compression strength (kN/m2)

1200

Long age strength estimation
In order to estimate the long age strength, thestrength increment (Δqu) from the initial strength(qu0) for the sand gel specimen of active colloidal
silica and silica sol was calculated using the latest test
data. The strength incremental rate (Δqu/Δt) obtainedby dividing the strength increment (Δqu) by theelapsed days (Δt) was shown in a vertical axis. The
curing temperature was shown in a horizontal axis asa reciprocal number of the absolute temperature (T).
As a result, the inclination of Arrhenius plot was
obtained as shown in Fig. 9. The acceleration
coefficient calculated from this Arrhenius plot was
multiplied by the sand gel age of heating curing. Thenthe standard curing strength was able to be predictedfrom the heating curing.

A

1000

B

800
600
400
200

0
1

10

100

1000

10000

100000Time (day)Fig.10Predictive strength of active colloidal silica
sand gel

Unconfined compression strength (kN/m2)

1200

The change of predictive strength for the active
colloidal silica sand gel is shown in Fig. 10, which
was calculated using the acceleration coefficientobtained from Arrhenius plot in Fig. 9. A measured
value was defined as the compression strength of theordinary temperature curing specimen. While a
predictive value was defined as the compression
strength which was derived from both theacceleration coefficient calculated from Arrhenius
plot and the material age of the heating curing
specimen. The measured values and predictive values
showed similar strength change in the specimens
whose material age was between 14 and 1, 000 days.
In addition, increase of compression strength wasable to be observed in both the measured value andthe predictive value of the specimens whose material
age was more than 60 days.

E

1000

F
G

800
600
400
200

0
1

Fig.11

10100Time (day)

1000

10000

Predictive strength of silica sol sand gel

From the results mentioned above, it was
confirmed that the predictive strength was able to be
derived from the heating curing. Table 3 shows theresults of the promotion testing in each case. From theresult, the strength of up to approximately 50 years
later at the longest was able to be predicted bymeasuring the strength of the specimen whose age
was 1, 000 days. That is, it was cleared that there wasa possibility of long-term strength prediction in ashort term.

The predictive strength of the silica sol sand gel
was shown in Fig. 11, which was obtained using theacceleration coefficient from Arrhenius plot in Fig. 9.
The results of measured values and predicted values
showed similar strength change, too.

178

GEOMATE- Bangkok, Nov. 14-16, 2016
Table 3.

Active colloidal silica

Forecast result

Test pieceCuringtemperaturesand gelplain gel

45℃

sand gel
Silica sol
plain gel

45℃
55℃
45℃
55℃

Acceleration Predictive time(year)
factor
17.84
48.9
17.3
6.30
9.2
3.35
16.5
6.01
8.9
3.25
11.8
4.32

CONCLUSIONREFERENCES

The efficiency of the promotion test with heating
curing was confirmed in all cases through theexperiments.
(1) It was proved that the active colloidal silica wasa permanent grout material without loss ofstrength with time elapse.
(2) The predictive value derived from Arrhenius plot
showed similar strength change to the measured
value.
(3) The prediction method for long-term strength of
improved soil by chemical injection in a shortterm was accomplished through the laboratory
experiment with a constant temperature tank.

[1] Sasaki, T., Oyama, T., Shimada, S., Suemasa,N(2012). “ Influence of Gelling Time On
Permeability And Strength of Ground Improved
By Chemical Grouting Method, ” The Twentysecond International Offshore and Polar
Engineering Conference, 17-22 June, Rhodes,
Greece[2] Sasaki, T (2013). “Liquefaction measures ofground seen in the case, ” Kindaikagakusha, pp
158-159.
[3] Nakayama, T., Sawada, R., Hiraoka, Y., Akagi,
H (2013). “ Evaluation of Long Term Stability ofChemically Stabilized Sand with Grouting. ”
Japan Socuety of Civil Engineering Memoirs C(Ground Zone Engineering), Vol69, No.2,
pp162-173.
[4] Geo-grouting development Organization HP:
http://www.jckk.jp/kokyu.php

179

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

VERIFICATION OF ELASTO-PLASTIC MODEL FOR CEMENTTREATED SOIL BEHAVIOR UNDER TRIAXIAL TENSIONCONDITION
Tsutomu Namikawa1
Department of Civil Engineering, Shibaura Institute of Technology, Japan

1

ABSTRACTThis paper presents the numerical simulation of the triaxial tension behavior of cement-treated soils. An elastoplastic model is adopted to simulate the triaxial tension behavior of the cement-treated soils. In the triaxial tension
test results, tensile failure is observed under low confining pressures and shear failure is observed under highconfining pressures. Thus two failure surfaces, tensile and shear failure surfaces, are assumed to express the two
failure modes observed in the laboratory tests. The simulation results agree reasonably with the laboratory test
results, indicating that the adopted elasto-plastic model can properly describe the tension and shear behaviors of
cement-treated soil under effective confining pressures.
Keywords: Cement-treated Soil, Triaxial Tension Behavior, Elasto-plastic Model, Numerical Simulation

INTRODUCTION

ELASTO-PLASTIC MODEL
A brief description of the elasto-plastic model
proposed by Namikawa and Mihira (2007) [1] isgiven here.
General Formulation
It is assumed that a total strain increment ij is asum of the elastic component ije and plastic
pcomponent ij , that is,

Ground improvement by cement mixing has been
widely used for structure foundation, retaining wall,liquefaction mitigation and so on. Recently column
and wall shaped ground improvements are often
adopted to reduce construction cost. When external
forces apply to the cement mixing column, tensilestress possibly occurs in the column under a low
confining stress condition. Thus the evaluation oftensile behavior of cement-treated soils is required todesign the ground improvement rationally. However
there is little study clarifying the tensile behavior of
cement-treated soil under a confining pressure.
The author has conducted the drained triaxial
tension tests of cement-treated soil to investigate thetension and shear failure behaviors under effective
confining pressures [1][2]. This paper presents thenumerical simulation of the triaxial tension test of
cement-treated soils. The elasto-plastic model
proposed by Namikawa and Mihira (2007) [1] isadopted to simulate the triaxial tension behavior of
the cement-treated soils. In the triaxial tension test
results, tensile failure is observed under low
confining pressures and shear failure is observed
under high confining pressures. Thus two failuresurfaces, tensile and shear failure surfaces, areassumed to express the two failure modes observed in
the laboratory tests. The simulation results agree
reasonably with the laboratory test results, indicatingthat the elasto-plastic model can properly describe thetension and shear behaviors of cement-treated soil
under effective confining pressures.

ij  ije  ijp(1)

The elastic strain increment is linearly related to aneffective stress increment  ij as1
ije  Eijkl
 kl

(2)

where Eijkl is the matrix of elastic constant. The stressstrain relationship of the elasto-plastic model can be
described as

f

E mnop op mn
Q ij  E ijkl  kl Ef
Q ijkl  kl
H
E qrst
 qr
 st
(3)
where f is the yield function, Q is the plastic potentialfunction and H is the plastic modulus.
Form of Failure and Yield Surface
In this model, two different failure surfaces are180

GEOMATE- Bangkok, Nov. 14-16, 2016

defined to describe the tensile and shear failures. The
failure function for the shear Fs is defined as

fs defined by Eq. (6). Then the internal variable kf isdefined as the following hyperbolic function of thedeviator plastic strain invariant : 

Fs  ij  J 2   c tan   p k f  0,

k y  k0

6 sin 2
kf 
,   ,
1     1    sin 3
3 3  sin  
3  sin ,   
3  sin 
66(4)

k f  k0
k0  p ey1 pey

1 sp sp
 ij  ij ,
2

, p

(10)

Tf

3  6c tan   p 

J 2 is the invariant of deviator stress, iswhere  ijsp is the deviator plastic strain tensor.  andthe friction angle, c is the cohesion coefficient, isthe Lode angle and p is the mean effective stress. It is
assumed that tensile failure occurs when a minor
principal stress reaches the tensile strength Tf. Thus
the tensile failure function Ft is described as

ey are the material parameters. Here it is assumed thatthe plastic volumetric strain does not occur in thestrain-hardening process. After the strain reaches theshear failure surface, the strain-softening behavior iscontrolled by the internal variable  defined aswhere  2 J 2 3 sin  23    p  T f  0

Ft  ij   1  exp Based on the failure surfaces, two different yieldsurfaces, the shear yield surface fs and the tensile yieldsurface ft, are introduced to describe the tension andshear strain-hardening/softening behaviors. The shear
and tensile yield surfaces are expressed in a form thatis similar to the failure surfaces. Then the yieldfunctions fs and ft are expressed asv p  Dc  p(7)
where ky is the internal variable denoting degree of
hardening and  is the internal variable denoting
degree of softening. While ky is used for the shear
yield surface to describe the strain-hardening
response,  is used in the strain-softening process. ky
and  are defined as  l c   T f
0  6 3

Tf
Gf  
P3p
3 peak if 0    0.75 (13)





Tf
12 1
 lc
 3P   3ppeak if 0.75    1
17 17 G f

where Gf is the fracture energy,  3p is the minor(8)pprincipal plastic strain and  3 peak is the minor

The elastic region is assumed to be defined as
f 0  ij  J 2 

(12)

(14)

  0, 0  k y  k f before Fs = 0 or Ft = 0
after Fs = 0 or Ft = 0

lc
(11)
t s0

where Dc is the dilatancy coefficient.
After the stress reaches the tensile failure surface,
the strain-softening behavior is controlled by . Herethe strain-softening behavior is defined by the bilinear
function. Thus,  is defined as2 sin     p  T f 1    0
3 

3

2 J2

0    1, k y  k f

Rl 

material parameter, ts0 is the thickness of shear bandand lc is the characteristics length related to the meshsize. In the strain-softening process, the plastic
volumetric strain is assumed to occur due to dilatancy.
The plastic volumetric increment v p is defined as

f t  ij ,  pwhere  peakis  p at peak stress state, er is the c1   
f s  ij , k y ,   J 2     p k y  0 tan 

(6)





 Rl  p   p peak 

,
e
r



(5)

principal plastic strain at the tensile peak stress state.
TRIAXIAL TENSION TEST

(9)

Namikawa and Hiyama (2014) have conducted
the drained triaxial tension and compression tests fora cement-treated soils [2]. A brief description of theexperimental procedure and result is described here.

Hardening and Softening Rules
In the strain-hardening process, the stress lies on
181

GEOMATE- Bangkok, Nov. 14-16, 2016

results of the triaxial tension tests. It can be seen that
all the samples show an approximately linear stressstrain relationship during the initial loading process.
For the sample tested at the value of pc’ = 100 kPa(AT-1), the deviator stress increased almost linearlywith the strain until the failure occurs and the strain at
the peak stress is around -0.015%. For the sample
tested at the value of pc’ = 500kPa (AT-5), although
the slight strain hardening is observed before the peakstress, the sample reaches the peak stress at small
strain level. Fig. 2 shows the results of the triaxial
compression tests. It can be seen that all the samples
show a strain hardening behavior before the peakstress. The absolute values of strain at the peak stressare around 0.7% and 1.1% for AC-1 and AC-3. For
Binder A, the absolute strain values at the peak stressin the triaxial compression tests are significant larger
than those in triaxial tension tests.
The typical stress-strain relationships for Binder
B are shown in Fig. 3 and Fig. 4. Fig. 3 shows theresults of the triaxial tension tests. It can be seen thatthe sample shows a failure at small strain level at thevalue of pc’= 30kPa (BT-1). This behavior is similar
to that observed in the result of Binder A. Conversely,
for the sample tested at higher values of pc’ (BT-6),
the strain hardening is observed before the peak stressand the absolute value of strain at peak stress becomes

Test Procedure
The material used in this study is a mixture of
Toyoura sand (D50 = 0.18 mm, Uc = 1.6), Portland
cement, water and Kaolin clay. The mixing
proportions are listed in Table 1. Two kinds ofbinders are prepared in the test. The specimen is madeby vibration after mixing thoroughly all the materials
that composed the cement-treated sand. The
cylindrical mold is used for preparing the specimens.
The curing time is 7 days. The unconfinedcompressive strengths qu are around 1300 kPa and500 kPa for Binder A and B. Before the shearing
process, the specimen is consolidated isotropically tothe prescribed effective confining pressures.
Thereafter the axial load is applied under a drained
condition. The experimental program is summarizedin Table 2.
Test Results
Stress-strain relationship
The typical stress-strain relationships for Binder
A are shown in Fig. 1 and Fig. 2. Fig. 1 shows the
Table 1 Mixing properties of cement-treated soil
Binder
A
B

Sand
69.3%
71.2%

Cement
7.0%
4.5%

Clay
5.0%
5.6%

Water
18.7%
18.7%
AT-1

Table 2 Summary of triaxial tests
AT-3
Case
AT-1
AT-2
AT-3
AT-4
AT-5
AC-1
AC-2
AC-3
AC-4
AC-5
AC-6
BT-1
BT-2
BT-3
BT-4
BT-5
BT-6
BC-1
BC-2
BC-3

Binder
A
A
A
A
A
A
A
A
A
A
A
B
B
B
B
B
B
B
B
B

pc’: kPa
100
200
300
400
500
50
50
100
100
150
200
30
100
200
300
400
500
50
100
200

Loading
TensionTensionTensionTensionTensionCompression
Compression
Compression
Compression
Compression
Compression
TensionTensionTensionTensionTensionTensionCompression
Compression
Compression

AT-5

Fig. 1 Stress-strain relationships of triaxial tension
test for Binder A.

AC-5
AC-3
AC-1

Fig. 2 Stress-strain relationships
compression test for Binder A.
182

oftriaxial

GEOMATE- Bangkok, Nov. 14-16, 2016

significantly larger than that for BT-1. Such
difference between the pre-peak behaviors may be
induced by the difference between the failure modes
depending on pc’. The failure modes under triaxial
tension condition is discussed in the following section.
Fig. 4 shows the results of the triaxial compression
tests. Similar to the result in Binder A, all the samples
show a strain hardening behavior before the peakstress. The absolute value of strain at the peak stressis around 1.8% for BC-1.
The triaxial tension and compression test results
show the difference of the stress-strain relationships.
For Binder A in which qu = 1300 kPa, the pre-peakstress-strain relationship is approximately linear
under the tiaxial tension condition while the strain
hardening behavior is observed under the triaxial
compression condition. For Binder B in which qu =500 kPa, the strain hardening behavior is observed
even under the triaxial tension condition. These
results indicate that the constitutive model for
cement-treated soil should describe such variation of
the stress-strain relationships under the triaxial
tension and compression conditions.

Failure surfaces
The peak stress states of all the cases are shown in
Fig. 5 and Fig. 6. In the results for BC-2 and BC-3,
since the peak stress state is not obtained clearly fromthe stress-strain relationships, the stress at the axial
strain of 2 % is defined as the peak stress. The shear
and tension failure surfaces defined by Eqs. (4) and(5) are plotted in these figures. The material
parameters of the failure surfaces are shown in Table
3. The values of c and which are required for the
shear failure surface are determined from the triaxial
compression test results. The value of Tf is determinedfrom the uniaxial tension test results. For Binder A,
the triaxial compression test results lie on the shear
failure surface defined by Eq. (4) while the triaxial
tension test results lie on the tension failure surface
defined by Eq. (5).
The difference of the failure modes may induce
the difference of the pre-peak stress-strain
relationships as mentioned in the previous section.

BT-1
BT-3
BT-6
AT-1
AT-5

Fig. 3 Stress-strain relationships of triaxial tension
test for Binder B.

Fig. 5 Peak stress state and failure surfaces for Binder
A.

BC-3
BC-2

BC-1

BT-1
BT-3
BT-6

Fig. 4 Stress-strain relationships
compression test for Binder B.

oftriaxial

Fig. 6 Peak stress state and failure surfaces for Binder
B.
183

GEOMATE- Bangkok, Nov. 14-16, 2016

For Binder B, the triaxial compression test results lie
on the shear failure surface. In the triaxial tension test
results, the peak stress states lie on the shear failuresurface under the high confining pressures while
those lie on the tension failure surface under low
confining pressures. It can be seen that the failuremode changes with the confining pressure for Binder
B. This difference of the failure modes in the triaxail
tension test results is consistent with the difference
between the stress-strain relationships in Fig. 3.

Simulation results
The comparison between the experimental andnumerical results for triaxial tension tests (AT-1, AT5, BT-1, BT-3, BT-6) are shown in Fig. 7 and Fig. 8.
Fig. 7(a) and Fig. 8(a) show that the numericalsimulation result reasonably agree with theexperimental result of tests AT-1 and BT-1 in which
the values of pc’ are low. This indicates the elastoplastic model can describe the tensile stress-strain
behavior under a low confining pressure. Fig. 7(b)and Fig. 8(b) shows that there is some difference
between the simulation and experimental results oftest AT-5 and BT-3. In these cases, it seems that theshear failure occurs in the simulation while thetension failure occurs in the laboratory test. Fig. 5 andFig. 6 show that the peak stress lies on the transition
region in those tests. These results indicate thatfurther research is required to simulate accurately thestress-strain relationship in the transition region. Fig.
8(c) shows that the numerical simulation result
reasonably agrees with the experimental result in test
BT-6. In this case, Fig. 6 shows that the shear failure
occurs in the test. It is clear that the elasto-plastic
model can simulate such shear hardening under thetiaxial tension condition.
The comparison between the experimental andnumerical results for triaxial compression test (AC-1,
BC-1) is shown in Fig 9 and Fig. 10. Althought there

SIMULATION OF TRIAXIAL TESTS
Simulation Conditions
The laboratory tests presented in the previous
section are simulated by the elasto-plastic model
proposed by Namikawa and Mihira (2007). The
elasto-plastic model is implemented as a user’s
subroutine in the finite element software DIANA.
The iterative initial stress method is used in solving
the problem incrementally. In the laboratory tests, thespecimens exhibit complex failure patterns after the
peak stress state and the global stress-strain
relationships depend on the failure patterns. It isdifficult to consider such complex failure pattern inthe analysis. Therefore the specimen is modeled as anelement and the predictions are restricted to the prepeak stress-strain relationships. Moreover one phaseformulation is used in the analysis by assuming theperfectly drained condition.
The material parameters used in the simulation are
shown in Table 3. Since the simulation is restricted tothe pre-peak regime, the parameters for the softening
behavior are omitted from this table. The values of theelastic modulus E are determined from theexperimental stress-strain relationship shown in theprevious section. As mentioned above, the values ofthe friction angle  and cohesion c are determinedfrom the peak stress states obtained from the triaxial
tests in this study. The values of the tensile strength
Tf are determined from the uniaxial tension test
results. The values of Poisson’s ratio , the hardening
parameters  and ey are determined based on theprevious test results [1].

(a)

(b)

Table 3 Material parameters
Parameter
E



c
Tf


er

Binder A
3000 MPa
0.167
33 degree
300 kPa
180 kPa
1.05
0.0002Binder B
1000 MPa
0.167
40 degree
90 kPa
50 kPa1.05
0.0002

Fig. 7 Comparison of stress-strain relations fortriaxial tension tests in Binder A: (a) AT-1 and (b)
AT-5.
184

GEOMATE- Bangkok, Nov. 14-16, 2016(a)Fig. 10 Comparison of stress-strain relation fortriaxial compression test (BT-1).

(b)

is some difference between the experimental andsimulated stress-strain relationships in the case ofBC-1), the simulation result agree reasonably with theexperimental result. This indicates that the elastoplastic model can also describe the triaxial
compression behavior of the cement-treated soil.
CONCLUSION
This paper presents the numerical simulation of
the triaxial tension test of cement-treated soils. The
elasto-plastic model proposed by Namikawa andMihira (2007) is adopted to simulate the triaxial
tension behavior of the cement-treated soils. The
simulated stress-strain relationships reasonably agreewith those obtained from the laboratory tests,
indicating that the elasto-plastic model can describe
the tension and shear behaviors under the effective
confining pressures. In particular, it is clear that the
adopted model can simulate the shear hardening
observed in the triaxial tension tests under higheffective confining pressures.

(c)Fig. 8 Comparison of stress-strain relations fortriaxial tension tests in Binder B: (a) BT-1, (b) BT-3
and (c) BT-6.

ACKNOWLEDGEMENTS
The author acknowledges the support of the Japan
Society for the Promotion of Science (JSPS
KAKENHI Grant No. 25420509).
REFERENCES[1] Namikawa T and Mihira S, “Elasto-plastic model
for cement-treated sand”,
Int. J. Numer.
Anal. Meth. Geomech., Vol. 31, No.1, 2007, pp.
71–107.
[2] Namikawa A and Hiyama S, “Triaxial tension
test of cement-treated soil”, Proceedings of the
15th Danube – European Conference on
Geotechnical Engineering, Vienna, Austria,
Vol. 1, 2014, pp. 329-334.

Fig. 9 Comparison of stress-strain relation for triaxial
compression test (AT-1).

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ASSESSMENT OF PHYSICAL AND MECHANICAL PROPERTIES OFCOMPOSITE TILE MADE OF WOOD CHIPS AND CERAMICPOWDER REINFORCED WITH BAMBOO FIBERS
Hermie M. del Pilar1, Marisol E. Manarin1, Clevin John V. Santos1,Krizza Diane M. Santos1,
1
Department of Civil Engineering, Malayan Colleges Laguna,Pulo Diezmo Road Cabuyao City, 4025, Philippines
ABSTRACT
Innovatıons in composite materials prompted this study to assess the physical and mechanical propertiesof three layer wood–ceramic composite tile. Specimens of 200 mm x 200 mm wood tiles having 2.5 mm top andbottom face layers and 5mm thickness core layer were produced to make up a 10 mm thickness of composite
tile. Face layers made of fine wood chips and ceramic powder mixture, having 70:30 weight ratio. Core layers
made of coarse wood chips binded with UF 12% by weight resin were prepared in 3 mix ratios with bamboo
fiber yielding ratios 80:20, 70:30, 60:40. Bamboo fibers reinforcement has a nominal diameter between 1.0 mm– 2.0 mm cross-oriented in one direction interfaced between face layers and core layers. The composite board
was hot-pressed 150OC at 1.38MPa in 5 minutes using Carver Laboratory Press of RI Chemicals Corporation.
The property of the boards were tested using Thickness Swelling (TS), Water Absorption (WA), Modulus of
Rupture (MOR) and Impact Strength (IS). The 60:40 ratio yields the highest MOR and IS tests at 37.55 MPa and83.42 Kpa-m respectively, while the 80:20 ratio is the lowest at 18.43 MPa and 59.49 KPa-m respectively. For
the physical property tests, there were no significant differences on TS having values close at 0%. WA resulted
to 8% - 9% for all mix ratios with no significant difference. This study concluded that the incorporation ofbamboo fibers will improve the mechanical properties in terms of MOR and IS.
Keywords: Composite Board, Bamboo Fibers, Urea Formaldehyde, Ceramic Powder, Bending Strength
polypropylene. Takagi and Ichihara [7] reported thatboth tensile and flexural strength of composite
manufactured boards with a mixture of starch- based
degradable resin and short bamboo fiber werestrongly affected by fiber ratio and fiber content.
Currently, the utilization of bamboo earned its placeon many researches for structural materials such as

INTRODUCTION
Manufactured composite boards arewidely used in the construction industry. The
attainment of optimum strength is one of theconcerns in the continuous development of the
products. One way to enhance the strength ofcomposite boards is to introduce a reinforcing
component.
Recently, researchers reported the great
potential use of bamboo for composite boards.
Bamboo is an ideal material for construction due toits characteristics, such as: fast growing, widelydistributed, excellent performance, and sustainable[1]. Bamboo is a porous material with axiallyarranged fibers that causes its mechanical andphysical properties to be directionally dependent [2].
Because of these properties, several research studies
were done using bamboo-based composite materials[3].
In the study of Chen, et.al[4].,a three-layercomposite board has improved its strength with theaid of bamboo- reinforced fibers, determined whenthe mechanical properties of bamboo fiber
reinforced polypropylene (PP) composite were
tested. Deshpande and Rao [5] also used extracted
bamboo fibers as reinforcement in polyester
composites.
Okubo,et.al [6] reported an
improvement in tensile strength and Young’s
Modulus of Composite made from bamboo fiber andcomposite boards.

This study aimed to determine theeffects of increasing weight ratio of the bamboo
fiber reinforcement on the physical and chemical
properties of composite tiles made of wood chips
and ceramic powder reinforced with bamboo fibers.
A comparison on the physical and mechanicalproperties of the produced composite tiles and thecommercially available wood tile was also
conducted.
METHODOLOGY
2.1 Raw Materials Preparation
Bamboo fiber reinforcement with anominal diameter between 1.0 mm – 2.0 mm weredrawn from commercially available raw sticks
product (Figure 1). These sticks are domestic
product usually with a nominal diameter of 0.5 cm
and length of 30 cm. With the aid of a thin knife, thebamboo sticks can be further reduced to the desired
nominal diameter of 1.0 – 2.0 mm fibers.
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GEOMATE- Bangkok, Nov. 14-16, 2016

The ceramic powder material used for the facelayer of the board was sourced from Mariwasa Siam
Ceramics Incorporated (Figure 2). The ceramic
powder is an inorganic non-metallic solid powdered,
which is the same material in manufacturing ceramic
tile products. Ceramic powder together with the finewood chips was mixed using a common adhesive.

Fig.1 Bamboo Fibers

Table 1
Board Manufacturing Criteria
Raw
Density

WeightMaterials
Content

Orientation
Ratio (%)

Resin

0.80g/cm3

Core Layer:
80:20
Cross
12 %
Bamboo Fibers
70:30
Section
60:40(%)

The target density and dimensions of theboard were used to compute for the total mass of thecomposite tile. Based on the total mass, 70% of themass constitute the face layers while 30% is coveredby the core layers and bamboo fibers were
manufactured based on the total mass. Likewise, theface layers had 70% fine wood chips and 30%
ceramic powder. The total mass of the core layer andbamboo fibers were varied according to three weightratios, 80:20, 70:30 and 60:40. Bamboo fibers wereoriented in a cross section manner interfaced between theFig. 2 Ceramics
PowderIn this study, Urea formaldehyde (UF) was usedas adhesive (Figure 3), manufactured and purchasedfrom RI Chemicals Corporation located in Pasig
City, Metro Manila, Philippines. This adhesive wasused to bind the wood chips (Figure 4), powderedceramic and bamboo reinforcement together (Figure
7).

face layers and core layer. Urea formaldehyde with a resin
content of 12% and 1% of the catalyst was used as abinder for the composite tile.

2.3 Machines for the composite board production
The machines used in the production of
three layer composite tiles were mixers and the heatpress. As shown in Figure 5, the Ross mixer from RI
Chemicals Corporation consisted of a double
planetary mixer equipped with classic rectangularstirrer blades mounted at the mix vessel. The
machine runs at a fixed speed set at 60 rpm for
mixing operations. Figure 6 shows the heat pressmachine used – the Carver Laboratory Press from RI
Chemicals Corporation. Each platen was aluminumwith 8 square inches size installed with a dial typetemperature gauge and manual temperature adjusted
to 150OC.

Wood chips collected from Southern Sky
Construction Supply Cabuyao Laguna were used as
raw material for the face and core layer of thecomposite boards. Fine wood chips were used forthe face layer base material and coarse wood chips
were used for core layer. Wood chips were ovendried at 105 Celsius for twenty-four hours (24h) andthen cured at room temperature for two weeks.

Fig. 3 Urea Formaldehyde

TargetFig 4. Wood Chips
Fig. 5. Ross Mixer

2.2 Board Production
Three-layer composite tiles reinforced
with bamboo fibers were manufactured.
The
manufacturing criteria of the boards are listed inTable 1.

Fig. 6 Carver Laboratory
Press

2.4 Board Manufacturing Process
After the materials and equipment
preparation, ceramics, fine and coarse wood chips,
urea formaldehyde and bamboo fibers were weighedusing a digital electronic balance for proper
proportioning. Figure 7 shows the ceramic powder,fine wood chips weighed together to form the facelayer composition placed in a Ross Mixer for mixing

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GEOMATE- Bangkok, Nov. 14-16, 2016

operation. Figure 8 shows the UF applied quickly to
avoid forming colloids due to the reaction of thecatalyst. The materials were mixed for five minutes
to allow homogeneous mix.

Fig.7. Face Layer
Composition in
the Ross Mixer

resulting boards were allowed to cure at room
temperature (Figure 12).

Fig. 11. Combined Layers Fig. 12. Hot-pressed
for Pressing
composite board

Fig. 8. Face Layer in
Mixed Proportion

2.5 Preparation of Test Specimen.

Figure 9 shows the mixture placed in form seats
prepared for the face layers of the board. The metal
mold has eight inches square in size and twelve
inches height. The surface of the face layer was
flattened using a heavy metal cover to spread the
mixture uniformly. In Figure 10, bamboo fibers wereplaced mutually perpendicular on face layer seats.
Urea formaldehyde was spread over the bamboo
fibers. Two layers of bamboo fibers in cross oriented
manner were formed. The bamboo fibers wereplaced slowly to avoid disruptions of the face layer.

Fig. 9. Bottom Face
Layer Seat

Specimens were prepared to determine thephysical and mechanical properties of the composite
tile product. In this study, 4 tests were conducted, 2for each property. The boards were tested for water
absorption and thickness swelling for physicalproperty tests, while the modulus of rupture andimpact strength account for the mechanical property
tests

Fig. 10. Bamboo Fibers
at the Face Layer

The wood chips and adhesive were mixed using
the Ross Mixer forming the core layer of the board.
The mixture was evenly spread on top of thebamboo fibers reinforcement. Another set ofbamboo fiber reinforcement was placed on top of thecore layer.

Fig. 17. Cutting pattern for test specimens testing

2.6.1 Physical Property Test
Water Absorption (ASTM D1037-99, 100-107)
The specimens with dimension 40mm x55mm are first weighed individually for this test,
then submerged to distilled water for the next 24hours, after which it was weighed again to determine
the water absorption of each specific sample. Water
absorption is defined as the increase in weightpercent and was determined using:

The top face layers were added with the same
composition of the bottom face layer. The top facelayer was the final part of the three-layer composite
tile. The three layers of the composite tile initiallypressed inside the mold using a heavy metal cover tocombine all 3 layers as one (Figure 11). The
composite tile was removed from the mold andplaced in the aluminum sheet for hot pressing.
The forming size of the board was 200 mm
wide and 200 mm long and the thickness wascalibrated at 10 mm thickness. The hot plate Carver
Laboratory Press temperature was adjusted to 150
degree Celsius with pressure of 1.38MPa, and the
hot press time was adjusted to 12 minutes.The

WA(%) =

Wet Weight − Dry Weight
Dry Weightx 100%(1)

Thickness Swelling (ASTM D1037-99, 100-107)
The specimen with dimensions 40mm x55mm were subjected for this test. First, thespecimen’s thickness was measured, and thensubmerged in distilled water for the next 24 hours.
Next, the samples were taken from the container and188

GEOMATE- Bangkok, Nov. 14-16, 2016

THICKNESS SWELLING TEST

TS (%) =Final Thickness − Initial Thickness
Initial Thickness

2.6.2 Mechanical Property Test

x 100%

Thickness Swelling (%)

the adsorbing water on surface was removed. The
thickness of the specimen was again measured todetermine the thickness swelling. Thickness
swelling is determined by the formula:(2)

Modulus of Rupture
A static bending test was conducted using
the Shimadzu, Universal Testing Machine. Three
specimens (190mm x 45mm x 12 mm) wereprepared for the static bending test using the 3 point
bending with effective span of 160 mm at a loading
speed of 5 mm/min. The Modulus of rupture can be
determined using the formula:
MOR =

3PL

2bd2

9.0
8.0
7.0
6.0
5.0
4.0
3.0
2.0
1.0
0.0

Commercial Wood Tile

80:20
70:30
60:40
Weight Ratio of Bamboo Fibers

Fig. 19. Thickness Swelling with Varying
Bamboo Fibers Weight Ratio
Water Absorption Test
Figure 20 shows the variations in water
absorptions measured for each weight ratio. The
water absorption of the composite tile was lesscompared to the commercial wood tile. The water
absorption increases with the increase in weight ofthe bamboo fiber. Yang, et.al. found that “Bamboo
fibers are highly hydrophilic due to their chemical
constituents such as lignin which can decreaseadhesion with hydrophobic matrix materials” [9],which is consistent with result.

(3)

Impact strength (ASTM A370)
The impact strength (IS) test measured theamount of energy or load required to break the board
specimens. The Charpy impact test, also known as
the Charpy V-notch test, is a standardized highstrain-rate test which determines the amount ofenergy absorbed by a material during fracture. Thisabsorbed energy is a measure of a given material's
notch toughness.

Water Absorption (%)

Water Absorption Capacity

2.7 Statistical Tool
The One Way Analysis of Variance(ANOVA) was utilized to determine if there is asignificant difference on the performance ofcomposite boards produced using the 3 mix ratios
for the physical tests (WA and TS) and the
Mechanical Tests (MOR and IS).

14.0
12.0
10.0
8.0
6.0
4.0
2.0
0.0

Commercial Wood Tile
8.045

8.343

8.837

80:20
70:30
60:40
Weight Ratio of Bamboo Fibers

Fig. 20. Water Absorption with Varying Bamboo
Fiber Weight Ratio

3. RESULTS AND DISCUSSION
3.1 Physical Property Tests Results

3.2 Mechanical Property Tests Results
Thickness Swelling Test
Modulus of Rupture
Figure 19 shows the changes in for each of theweight ratios.
The thickness swelling of thecomposite tile is less than the commercial wood tile.
The results showed that the thickness swellingincreases with the increase on bamboo fiber content.

Figure 21 shows the results of Modulus
of Rupture for each weight ratio. The means ofModulus of Rupture increases as the weight ratio
increased. The increasing Modulus of Rupture
affirms the claim of Hermawan et.al (2011) relating
the uniform dispersion of the fibers in the face layerof the board which leads to the even distribution andtransfer of stress from the face layer to the fibers[10].

“Bamboo is able to absorb all fluids of different
densities over the time most likely due to thecapillary forces within the pore structure” [8], whichis consistent of the result.

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Modulus of Rupture (MPa)GEOMATE- Bangkok, Nov. 14-16, 2016

50.0
45.0
40.0
35.0
30.0
25.0
20.0
15.0
10.0
5.0
0.0

Mechanical Strength (MOR)
Commercial Wood Tile

Table 2: Test for the difference in properties of thecomposite tile37.55928.213Comparison
Between

18.934

3 Types ofbambooreinforced
tile board

80:20
70:30
60:40
Weight Ratio of Bamboo Fibers

Impact Strength (KPa-m)

Signific
ant

IS

WA

TS

SignificantNot
Significant

Not
Significant

In terms of the physical properties, theproduced composite board had lower values for boththickness swelling and water absorption. For
mechanical properties, the board exhibited excellent
results with the variation of weight ratio of thebamboo fibers.
Based on the outcomes, several conclusions weredrawn.

Figure 22 shows the result recorded from
Charpy Impact Strength Test on nine specimens foreach weight ratio. The means of the impact strength
increase as the weight ratio of the bamboo fibers
increased. This shows that the reinforcement
provided by the bamboo fibers results to asignificant increase in the impact strength of thecomposite tile. “The increased fiber content is
improving its capability to absorb impact, which isconsistent with the findings of Jena and Pradhan[11].
Impact Strength Test

MOR

Physical
Property

4. CONCLUSIONS

Fig. 21. Modulus of Rupture with Varying
Bamboo Fibers Weight Ratio

80.0
70.0
60.0
50.0
40.0
30.0
20.0
10.0
0.0

Mechanical
Property

59.48662.20265.609

80:20

70:30

60:40

1.) The water absorption and thickness swelling of
the composite board had shown good quality
regardless of the weight ratio used. The
thickness swelling was almost 0%. The low
values of water absorption and thickness
swelling implied that the boards have highresistance to moisture content.
2.) The bending strength of the board issignificantly affected by weight ratio of bamboo
fibers. The Modulus of Rupture shows larger
value with increased weight ratio of bamboo
fibers. Among the weight ratios 60:40, 70:30
and 80:20, the best value of MOR is 37.55 MPa
from the 60:40 ratio, while the lowest value is
18.43 MPa from 80:20 ratio.
3.) The Impact Strength increased as the weightratio of the bamboo fibers increased. The 60:40
ratio exhibits the highest value of 65.609 KPa-m
while the ratio 80:20 exhibits the least value of
59.49 KPa-m. The increased fiber content is
improving its capability to absorb more energy.
This shows that the reinforcement provided by
the bamboo fibers results to a significant
increase in the impact strength of the composite
tile. Impact strength of the composite tile withbamboo reinforcement exhibited excellent
quality when compared with commercial wood
tile whose impact strength measures 35.47 KPam on average.
4.) The physical appearance and property of thecomposite board are comparable to the qualityof a commercial wood tile due to its surface
glaze and rigidity. The tapestry brought by thecompressed ceramic powder and fine woodchips at the face layer significantly add to theaesthetic value of the composite board.

Weight Ratio of Bamboo Fibers

Fig.22 Impact Strength with Varying Bamboo
Fibers Weight Ratio

Statistical Results
The statistical results using One-Way
ANOVA showed that the use bamboo strips
significantly affect the Mechanical Property but not
the Physical Property. This is using an α = 0.05.
Table 2 summarizes the result for the test ofsignificance.

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GEOMATE- Bangkok, Nov. 14-16, 2016

REFERENCES:[6] Okubo, K., Fujii, T., Yamamoto, Y. (2004)
Development of Bamboo-Based Polymer 59 ,Composites and their mechanical properties.
Composite: Part A 35:377 – 83

[1] Yao, W. & Zhang, W. (2011). Study on Bamboo
Conditions for Processing Bamboo Fiber.
Fourth International Conference on Intelligent
Computation Technology and Automation.
10.1109, 198-202. 62[7] Takagi, H., Ichihara, Y. (2004). Effect of fiber
length on mechanical properties of green composite
using a starch based resin and short bamboo fibers.
JSMEIntJ : Series A 47:551-5[2] Prieto, M.G., Vidal, N.J., & Rojas, J.T., (2013).
Bamboo as a Sustainable Material for
Several City Uses.Fourth International Conferenceon Intelligent Computation
Technology andAutomation. 978-1-4799-2911-5/13.

[8] Sobrinho, A., Aragao, J.L., Torres, S.M., Barros,
S., Ortiz, S.R. & Barbosa, N.P. (2009). Bamboo PH
and Absorption in Different Liquids.
[9] Yang, H., Kim,H., Park, H., Lee,B., & Hwang, T.
(2006). Water Absorption Behavior and Mechanicalproperties of Lignocellulosic filler-polyolefin biocomposites. Composite Structures. Vol. 72, no. 4 ,pp.429-437, 2006

[3] Suhaily, S., Khalil, A., Nadirah, W., & Jawaid,
M. (2013). Bamboo Based Biocomposites Material,
Design and Applications.Creative Commons
Attribution License. 10.5772/56057.

[10] Hermawan,A., Ohuchi,T., & Fujimoto, N.
(2010). Manufacture of Three-Layer Wood
Porcelain Stone Composite Board Reinforced with
Bamboo Fiber. Materials and Design. 32, 2485-2489[4] Chen X., Guo Q., &Mi Y. (1998) Bamboo FiberReinforced Polypropylene Composites: A Study of
the Mechanical Properties.JApplPolym Sci.69:1891–
1899.

[11] Jena, H., Pandit, M., & Pradhan, A. (2012).
Study the Impact Property of Laminated Bamboo
Fibre Composite Filled with Cenosphere.
International Journal of Environmental Science andDevelopment, Vol. 3, No. 5. 457-458.

[5] Deshpande A., Rao L., & Rao B. (2000).
Extraction of Bamboo Fibers and Their Use as
Reinforcement in Polymeric Composites.
JApplPolym Sci. 76:83–92.

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

UNCONFINED COMPRESSIVE STRENGTH OF COMPACTED DISTURBED CEMENT-STABILIZED SOFT CLAY
1

Mohamed Ayeldeen1, Yuki Hara2 and Masaki Kitazume3
Environmental Engineering Department, Egypt–Japan University of Science and Technology (E-JUST),
Alexandria, Egypt.
2, 3Civil and Environmental Engineering Department, Tokyo Institute of Technology, Tokyo, Japan.

ABSTRACT
Pneumatic flow mixing method is a new land reclamation method, developed in Japan to meet the persistentlack of space. In this method dredged soft soil is mixed with a small amount of stabilizing material (such as cement) during transporting the soft soil in a pipe using compressed air to be used for land reclamation. In somecases, the soil/cement mixture is stored in temporary place for days and then transported and compacted at therequired place. Basically, the cement chemical reaction starts immediately after the mixing with the soft soil andthe mixture starts to gain its strength, therefore disturbing the mixture after days from the mixing influences the
mixture strength. However, the soil/cement mixture is still able to gain extra strength after disturbance, transportation, and compaction. This study aims to evaluate the effect of dynamic compaction on the shear strength of
disturbed cemented soft soil mixture experimentally. The mixture was fully disturbed after one week from mixing with cement. Three cement/soil ratios were used in this study under different dynamic compaction energies.
Unconfined compression test was conducted at various curing times for both disturbed and non-disturbed specimens.
Keywords: Pneumatic flow mixing, Stabilized soft clay, Compaction, Unconfined compressive strength, Soil disturbancethis reformed connection will be. In this study the
disturbed period has chosen to be as long as possible,
to represent optimum reduction of the soil strength
after disturbance, where the chosen period is 7 days
from mixing the paste till the disturbance and themolding.
This study is an attempt to understand and evaluate the unconfined compressive strength of compacted disturbed cement-stabilized soft clay. Threecement/clay ratios were used in this study (5, 10 and15%). Different compaction energies were used tocompact the disturbed soil, while tapping was used
for non-disturbed soil. Unconfined compression test
was performed for both disturbed and non-disturbed
specimens after different curing times.

INTRODUCTION
Soft soil deposits can be found in many places allover the world especially coastal areas, which arerecognized by the high compressibility and low
shear strength. Where it became essential to findappropriate method to improve the softy soil characterizations to meet the needed engineering requirements for construction. One of the new methods
used to improve the soft soil for land reclamation ispneumatic flow mixing method, which depends on
mixing the dredged soft soil with cement duringtransporting the soil through pipes to the construction site [1], [2], [3], [4]. However, in some cases thesoil/cement mixture has to be stored in other placefor days and then transported again to the construction site. In that case the soil/cement mixture become disturbed, where the cement chemical reaction
already started as soon as the cement has been mixed
with the soil. The disturbance process, which is necessary to transport the mixture after days to the site,
shall change the mixture behaviour especially afterusing compaction to place the mixture in its final site.
Hence the chemical reaction of the cement already started once the cement has been mixed withthe soil, the efficiency of the connections reformed
after the disturbance is affected by the time period
spent from mixing paste till the disturbance and setting up again. The longer this period is, the weaker

MATERIALS AND TEST PROCEDURE
Material properties and specimen preparation
In order to have a clear understanding of the effect of compaction on the mechanical properties ofdisturbed cement-stabilized soft clay, Kaolin clay
(w L of 77.5% and w P of 30.3 %) and ordinary Portland cement were used in this study.
Kaolin clay was mixed with tap water for one
hour in a vacuum mixer at a water content ratio of
120%. Ordinary Portland cement was added after tothe slurry at cement content varying from 5 to 15%,192

GEOMATE- Bangkok, Nov. 14-16, 2016

does not exceed 50 kPa regardless the compactionenergy, C.E, and with failure stain of 6-7 % in a ductile behaviour with unclear peak strength.

and both clay slurry and ordinary Portland cement
were mixed for 10 mins. Cement content, aw, can be
defined as the ratio between the weight of cement
and the weight of clay in dry case. The paste wasthen split in two parts (one part for non-disturbed
specimens and the other for disturbed specimens).
For non-disturbed specimens, the paste was used
immediately after mixing and molded by tapping ina plastic mold, 50 mm diameter and 100 mm heightat three layers [5]. Each layer was tapped 100 times
to grantee the homogeneity of the specimens. The
non-disturbed specimens were kept in humidity curing chamber and were tested after 7, 28, 35 and 42days from mixing the paste.
For disturbed specimens, the paste was kept airtight in plastic bag for 7 days before it fully disturbed for 5 mins in the mechanical mixer. Specimens were molded in the same plastic molds as in
the non-disturbed case, however various dynamic
compaction energies were used to mold the disturbed specimens. The molds were filled at threelayers; each layer was dynamically compacted by 1
kg cylinder hummer falling free from a height of 1
meter. Different numbers of blows (5, 10, 20 and 40)
were used to compact each layer to obtain different
compaction energies. The tapping technique wasalso used to compact disturbed specimens at compaction energy, C.E. of zero. The specimens were
kept also in humidity curing chamber and were tested after 28, 35 and 42 days from the initial mixing.

Unconfined compressive strength,qun (kPa)

300
250
200
Non-disturbed

150
C.E.= 1840
kN.m/m3
C.E.= 920 kN.m/m3

100
50
0
0

Fig. 1

5

10
Axial Strain (%)

15

20

Stress strain curve for disturbed and nondisturbed specimen (aw = 15%, 28 curingdays).

The effect of compaction energy on the behaviour of disturbed cemented clay can be observed in
Figure 2 for the same conditions as in Figure 1 (aw =15%, t = 28 days). The ductile behaviour for disturbed cemented clay changed to be more brittle
with increasing the compaction energy. It can benoticed that the unconfined compressive strength
increased with increasing the compaction energy,
while the axial strain at failure decreased. For noncompacted specimen, the peak unconfined compressive strength, q un was about 21 kPa at axial strain at
failure of 12%. However, the peak strength increased to 24 kPa with increasing the compactionenergy to 460 kN.m/m3, where also the axial strain
at failure reduced to 8%, which reflects increasing
the brittle behaviour with increasing the compactionenergy. The brittle behaviour became more clear for
specimen compacted with higher dynamic energy, atcompaction energy of 1840 kN.m/m3, the peak unconfined compressive strength reached 28 kPa withaxial strain at failure of 6%.

Test procedure
Unconfined compressive strengths were obtainedusing a strain rate of approximately 1 %/min for all
specimens. Corrections to the cross-sectional area
were made prior to calculating the compressive
stress on the specimen. All tests were accomplishedthree times at least for each case. Curing time was afundamental parameter for understanding the behavior of compacted disturbed cemented-soft clay;
therefore, experimental schedule was arranged toperform the tests at different time periods (7, 28, 35,
and 42 days) from mixing the clay with the cement.

35
Unconfined compressive strength,qun (kPa)

30

RESULTS AND DISCUSSION

25

Effect of disturbance on the stress strain behaviour

20
15
C.E. = 1840 kN.m/m3
C.E. = 920 kN.m/m3
C.E. = 460 kN.m/m3
C.E. = 230 kN.m/m3
C.E. = 0.00 kN.m/m3

10

Figure 1 shows the difference in behaviour between disturbed and non-disturbed cemented clay
for cement ratio, aw of 15% tested after curing period, t of 28 days. For non-disturbed specimen, a clear
brittle behaviour with peak strength and quite small
residual strength can be notice with unconfinedcompressive strength, q un_Non_28 of 270 kPa at axial
strain at failure of 2%, while the unconfined compressive strength for disturbed specimen, q un_D_28

500Fig. 2193

5

10
Axial Strain (%)

15

20

Stress strain curve for disturbed specimen
with different compaction energies (aw =15%, 28 curing days).

GEOMATE- Bangkok, Nov. 14-16, 2016

For more understanding to the effect of compaction energy and curing time on the disturbed cemented clay, Figure 4 displays the effect of compaction energy on modulus of deformation, E 50 for aw =15% with different curing periods. It can be seenfrom the figure the increasing in modulus of deformation, E 50 with increasing the compaction energy.
At curing time of 28 days, E 50 has increased from1000 to 1250 kPa by increasing the compaction energy from zero to 920 kN.m/m3 and it has increased
again to be 1500 kPa at compaction energy of 1840
kN.m/m3 with increasing percentage of 125 and
150 % for both compaction energies of 920 and1840 kN.m/m3 respectively.

Effect of curing time on both unconfined compressive and modulus of deformation
Figure 3 clarifies the variation of q un under fivedifferent compaction energies with time for cement
content of 15%. Increasing the compaction energy
reflected to a positive increasing in the unconfinedcompressive strength. At curing time of 28 days, q unincreased from 25 kPa at zero compaction energy toabout 35 kPa at compaction energy of 1840
kN.m/m3 (i.e. the percentage of increasing is about140%). The same percentage of increasing can bealso notice at curing time of 42 days, where q un increased from 32 kPa to 46 kPa (increasing percentage of about 143%) by increasing the compactionenergy from zero to1840 kN.m/m3. It can be observed also that doubling the compaction energy
from 920 to 1840 kN.m/m3 did not have the considerable effect on the compressive strength, where q un
at curing period of 28 days had increased from 33 to
35 kPa after increasing the compaction energy from
920 to 1840 kN.m/m3. The same trend can be noticeat curing time of 42 days, where increasing the compaction energy from 920 to 1840 kN.m/m3 effected
on increasing the compressive strength from 43 to
46 kPa, i.e. increasing the compaction energy from
920 to 1840 kN.m/m3 caused an increasing in q unwith a percentage of 5-6 %. The unconfined compressive strength improved by percentages of 135,
and 142 % by increasing the compaction energies
from zero to 920 and from zero to 1840 kN.m/m3
respectively. This can be interpreted due to the highwater content and the very low permeability of clay
which affected the efficiency of compaction after
certain compaction energy.

Modulus of deformation, E50, (kPa)

2500

2000

1500

1000
C.E. = 0.00 kN.m/m3
C.E. = 230 kN.m/m3
C.E. = 460 kN.m/m3
C.E. = 920 kN.m/m3
C.E. = 1840 kN.m/m3

500

0
0

Fig. 410

20
30
Time,t (days)

40

50

Variation of modulus of deformation withtime under different compaction energies(aw = 15%).

50

The same trend still exists at curing period of 42days, where E 50 has augmented comparing the zerocompaction case with increasing percentage of 123
and 142 % for both compaction energies of 920 and1840 kN.m/m3 respectively. Unlike unconfinedcompressive strength, the effect of doubling thecompaction energy are more effective on E 50 thanq un , where E 50 has increased from 123-125% to 142150% by increasing the compaction energy from 920
and 1840 kN.m/m3.

Unconfined compressive strength, qun (kPa)

45
40
35
30
25
20
C.E. = 0.00 kN.m/m3
C.E. = 230 kN.m/m3
C.E. = 460 kN.m/m3
C.E. = 920 kN.m/m3
C.E. = 1840 kN.m/m3

15
10
5

Relationship between unconfined compressive
strength and modulus of deformation for both
disturbed and non-disturbed specimens

00Fig. 310

20
30
Time,t (days)

40

50

According to previous studies [1] [6] for undisturbed cemented Japanese clays and Bangkok clay,
It was found that E 50 can be taken as approximate
value of 50 to 300 × q un . However, in this study thecorrelation value of E 50 can be considered as 160 q un
as it can be seen in Figure 5 (a). This difference between the current study and the previous studies can

Variation of unconfined compressive
strength with time under different compaction energies (aw = 15%).

194

GEOMATE- Bangkok, Nov. 14-16, 2016

be explained due to the difference in used clay andwater content, however both values can be acceptable to estimate E 50 from the corresponding q un .
Figure 5 (b) shows the relation between E 50 andthe corresponding q un for disturbed compacted cemented clay with different cement ratios and different compaction energies. As mentioned before in
Figure 1, that the disturbance process reduced theunconfined compressive strength of the cemented
clay and transferred the behaviour to be ductile, thisbehaviour can be seen clearly form the approximate
function between E 50 and the corresponding q un . The
multiplier factor to estimate E 50 has considerably
reduced from (E 50 =160 q un ) for non-disturbancecemented clay to (E 50_D = 25 - 50 q un_D ) for disturbance cemented clay.

ing in shear strength, and increasing in the bearing
capacity. From Figure 6, increasing the compactionenergy affected positively on dry density, where theaverage dry density - at aw = 10% - has increased
from 7.0 to 7.3 kN/m3 by increasing the compactionenergy from zero to 460 kN.m/m3, and has reached
7.5 kN/m3 at compaction energy level of 920
kN.m/m3. Increasing dry density with increasing thecompaction energy was expected due to the effect ofcompaction on reducing void space between particles, which leaded to increasing in the density.
However, the compaction efficiency can affect bysome factors such as high water content, where it
can be notice at low level of compaction energy.
Increasing the compaction energy from zero to
230 kN.m/m3 did not have the noticeable effect on
increasing the dry density, while this effect started tobe obvious at starting from compaction energy of
460 kN.m/m3. Cement content affected the dry density, where increasing the cement content was followed by raising in the dry density. Dry density increased from 6.7 to 7.5 kN/m3 by increasing aw
from 5 to 10%, and to 8.1 kN/m3 at aw of 15%. Thiscan be explained due to increasing the cemented
pozzolanic formations inside the soil gaps, which
increased by increasing the cement ratio.

Modulus of deformation, E50, (MPa)(a) 7060
E50 = 160 qun

50
40
30

E50 = 147 qun
Lorenzo & Bergado(2006)

20
10
0
0

100

200

300

400

Unconfined compressive strength, qun (kPa)

Modulus of deformation, E50, (kPa)

(b)

3000
2500

E50_D = 50 qun_D

2000
1500
1000

Fig. 6500

E50_D = 25 qun_D

Relationship between unconfined compressive
strength and both dry density and voids ratio

0
0

10

20

30

40

50

Unconfined compressive strength, qun (kPa)

Fig. 5

Effect of compaction energy on dry density

Figure 7 displays the relationship between dry
density and disturbed unconfined compressive
strength for different cement ratios. This relationshipwas effected by changing the cement content, andthis can be notice clearly from the slope of the threetrend lines (A, B and C). The slope of trend line A
with the horizontal projection is almost neglectable,
which means that changing the dry density does nothave considerable effect on q un for aw = 5%, where
the dry density changed from 6.3 to 6.9 kN/m3 withaverage change in q un was from 6.5 to 8.5 kPa. The
slope of trend line B is more apparent than the trend

Modulus of deformation versus unconfinedcompressive strength of: (a) non-disturbed
cemented clay; (b) disturbed cemented clay

Effect of compaction energy on dry density
The plot of dry density against compaction energy for various cement ratios is shown in Figure 6,
where increasing the soil density is the key target forcompaction process. Increasing the soil density willlead to reduction in the ground settlement, increas195

GEOMATE- Bangkok, Nov. 14-16, 2016

unit weight, and G s = the mixture specific gravity.
The formation of e/aw has already collected theeffect of several variables such as water content,
curing time, cement ratio, and compaction energy.
The reason behind the division of e by aw that somespecimens can have different cement ratios andcompaction energies, however it may chance to havethe same void ratio, and for sure not the same valueof q un_D . The following empirical relationship has
been delivered to estimate the value of q un_D as clear
in Figure 8:

line A, which clarifies the relationship between dry
density and q un_D for aw = 10%, where the average
dry density has changed from 6.9 to 7.7 kN/m3 accompanied by changing in q un_D from 15 to 27 kPa.
That relationship reached the utmost clarity in caseof trend line C (aw = 15%), where the changing indry density – just from 7.8 to 8.2 kN/m3 - was followed by high changing in q un_D – from 26 to morethan 40 kPa.

qun_D = 700 × e

(2)

aw

Estimating the value of the increasing factor iFig. 7

It is important to find a strategy to estimate both
unconfined compressive strength and modulus ofdeformation for disturbed cemented clay under different compaction energies for various curing periods. A clear relationship between unconfined compressive strength for non-disturbed specimens at
curing time t , q un_Non_t and unconfined compressive
strength for disturbed non compacted specimens at
the same curing time t, q un_D_T_t can be figured out
from Figure 9, where it can be written as following:

Relationship between dry density and disturbed unconfined compressive strength.

qun_D_T_t = 0.1 × qun_Non_t

where both q un_Non_t and q un_D_T_t have the same
cement ratio, water content and curing time.

35
30

40

25
20

35

15

30

10
5
0
0

20

40

60

80

e / aw

Fig. 8

25
20
15
10
5

Unconfined compressive strength versus
e/aw.

0
0

The final void ratio corresponding to each dry
density - which takes into consideration the effect ofwater content, curing time, and compaction energycan be calculated as follows:

e=

(3)

40

qun_D_T_t (kPa)

Unconfined compressive strength, qun (kPa)45(1 + wc ) × Gs × γ w − 1
γtFig. 9

(1)

100

200
qun_Non_t (kPa)

300

400

Unconfined compressive strength for nondisturbed specimens versus unconfinedcompressive strength for disturbed noncompacted specimens

For disturbed cemented clay, the unconfinedcompressive strength will increase than q un_D withincreasing the compaction energy as mentioned be-

where γ t = the total unit weight after curing, wc =the water content corresponding to γ t , γ w = water
196

GEOMATE- Bangkok, Nov. 14-16, 2016

Obtaining a formula for the increasing factor ishall be helpful to estimate the value of unconfinedcompressive strength for disturbed cement clay for
different cement ratio, compaction energy, and curing time.

fore in Figure 2. This increasing can be represented
by the factor i, as following:(4)
q un_D_comp_t = i × q un_D_28
where, q un_D_comp_t is the unconfined compressive
strength for disturbed cemented clay under compaction energy, after curing period of t. q un_D_T_28 is theunconfined compressive strength for disturbed cemented non compacted clay after 28 days curing
period for the same cement ratio, and i the increasing factor as percentage.
It was not possible to figure out a formula to estimate i statistically, thus using neural network tool
will be essential to estimate the value of i as a function of both curing time and compaction energy,
which will be discussed in future.

ACKNOWLEDGEMENTS
The first author would like to thank Egyptian
Ministry of Higher Education (MoHE) for providing
the financial support (PhD scholarship) for this research as well as the Egypt–Japan University of Science and Technology (E-JUST) for offering the facility and tools needed to conduct this work. The
first author wishes to express his gratitude to Tokyo
Institute of Technology for providing the neededmaterials and apparatuses where this work would nothave been possible without that valuable support.

CONCLUSIONSREFERENCES

This study focuses on investigating the behavior
compacted disturbed cemented soft clay under different compaction energies and curing periods.
Three different cement contents are used in thisstudy with water content of about 120%. Within theexperimental range of this investigation, the following conclusions can be drawn:
Disturbing the soil has a serious effect on the
stress strain behaviour, where the behaviour changes
from being brittle with peak to more ductile, with ahuge reduction in the compressive strength andmodulus of deformation.
The stress strain behavior has changed slightly
due to the compaction, where increasing the compaction energy changed the behaviour to be morebrittle with increasing both unconfined compressive
strength, q un and modulus of deformation, E 50 . The
relationship between q un and E 50 for undisturbed
cemented clay has an acceptable matching with theprevious test results.
Increasing the compaction energy reflects to increase the soil dry density, however the compaction
mechanism can affect by some factors such as highwater content and soil particles size especially at low
levels of compaction energy.
The relationship between unconfined compressive strength, q un and both dry density and voids
ratio affects by many factors such as the cement
ratio, water content, and the compaction energy,
therefore using the term e/aw can be more reliable torelate q un with the soil void ratio e.

[1] Ministry of Transport, The Fifth District Port
Construction Bureaus, "Pneumatic flow mixing
method", Yasuki Publishers, 157p. (in Japanese).
[2] Kitazume, M., & Hayano, K., " Strength properties and variance of cement-treated ground using the pneumatic flow mixing method", Proc.
Institution of Civil Engineers - Ground Improvement, 2007, Vol. 11, No. 1, pp. 21–26.
[3] Kitazume, M. & Satoh, T., "Development of apneumatic flow mixing method and its application to Central Japan International Airport construction". Proc. Institution of Civil Engineers Ground Improvement, 2003, Vol. 7, No. 3, pp.
139–148.
[4] Kitazume, M. & Satoh, T., "Quality control inCentral Japan International Airport construction", Proc. Institution of Civil Engineers Ground Improvement, 2005, Vol. 9, No. 2, pp.
59–66.
[5] Japanese Geotechnical Society, "Practice formaking and curing stabilized soil specimens
without compaction", JGS 0821-2009. Japanese
Geotechnical Society. 2009, Vol. 1, pp. 426-434(in Japanese).
[6] Lorenzo, G. A., & Bergado, D. T.,"Fundamental
characteristics of cement-admixed clay in deep
mixing", Journal of Materials in Civil
Engineering, 2006, Vol. 18, No. 2, pp. 161–174.

197

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

A MODIFIED HARDENING SOIL MODEL FOR ROCKFILL MATERIALS
R. Sukkarak1, and P. Jongpradist1
1

King Mongkut’s University of Technology Thonburi, Bangkok, Thailand
ABSTRACTThe hardening soil (HS) model is widely utilized for simulating the stress-strain behavior for different soil
type, because many essential features of rockfills have been included in the hardening soil model (i.e.,
nonlinearity, stress-dependent stiffness and shear dilatancy). More recently, this model is adopted to characterize
the behavior of rockfill materials. However, several significant features of rockfill material have not included in
the current model, since the model was originally developed for fine grained soils (i.e., sands and clays). In thispaper, an existing HS model is modified for simulating the stress-strain-volumetric strain response of rockfill
material. A new hardening parameter, which is the stress-dependent stiffness value, is derived in order toimprove the simulation capability. A modification of Rowe’s stress-dilatancy theory is proposed to account for
the influence of particle breakage. The modified model predicts in very consistent manner with the triaxial
response of different type of rockfill materials.
Keywords: Hardening soil model, Rockfill materials, Stress-dependent stiffness value, Rowe’s stress-dilatancy
theory, Particle breakage

give a good agreement with experimental result or
dam monitoring data. However, some of model
parameters are often obtained by the backcalculation method or can be difficult to detemine
due to lacking of experimental result (i.e. oedometertest). Although many essential features of rockfills
have been included in the HS model (i.e.,
nonlinearity, stress-dependent stiffness and shear
dilatancy), the model was originally developed forsoils (i.e., sands and clays) [12]. Based on
experimental studies [5], [13], these studies
indicated that the particle breakage happen even low
confining pressure condition and changes in theparticle during shearing. This feature may the resultin different behavior between rockfill material andfine grained soils.
In this study, the HS model is developed for
simulating the stress-strain-volumetric strain

1. INTRODUCTION
Rockfill material is widely used in many
geotechnical applications, especially rockfill dams.
Based on several experimental studies [1]-[5] it has
widely been recognized that (1) the stress-strain
behavior of rockfill materials is non-linear and stressdependent, (2) the behavior of the rockfill materials
is affected by many factors as mineralogicalcomposition, particles grading, size and shape ofparticles, (3) the confining pressure is significant
that affects to stress-strain behavior of rockfill
materials, increase of confining pressure have an
influence to increase in deviator stress, axial strain
and volumetric strain at failure, (4) the shear
dilatancy phenomenon occurs when confiningpressure is low. For simulating these behaviors, thepreferred constitutive models must be based on anelasto-plasticity theory [6], intense shear dilatancy
behaviors of dense granular materials and stressdependence of stiffness.
The Hardening Soil (HS) model is the one of thekind of an elasto-plasticity model which isextensively used for simulating the stress-strain
behavior of many soil types. Several researchers
have reported that the HS model can be applied innumerical simulations of rockfill behavior bycomparing with the experimental data [7]-[8], andimplemented in rockfill dam stress-deformation
analyses [9]-[11]. Since, the HS model simulations( q − ε 1 − ε v ) response of rockfill materials. The
comprehensive testing results of Nam Ngum 2
rockfill materials, which have been well documented(including triaxial and oedometer tests), are used toinvestigate the significance of the effects of thestress-dependent stiffness and particle breakage. A
modification to Rowe’s dilatancy equation is
introduced in the current work to account for particle
breakage. The capability of the model has beenassessed by comparing between the predicted results
and test data.

198

GEOMATE- Bangkok, Nov. 14-16, 2016

(Tangent modulus) tests, ϕ is the friction angle, c

2. HARDENING SOIL MODEL

is the cohesion, p ref is the reference confining

The Hardening Soil (HS) model has been adopted tocharacterize the behaviour of different kinds of soil.
The HS is the developed constitutive model which
has been implemented in the finite element code
PLAXIS. The constitutive model associated with theelasto-plastic theory by introducing two yieldsurfaces as shown in Fig. 1(a). First, the Mohr–
Coulomb yield surface f has been specified in thepressure and m is the defined as the power for the
stiffness dependent.

following form:
f= f − γ

f
=

p

(1)
q

2

Ei 1 − q / q a2q−γ = − (2ε − ε ) ≈ − 2ε
p

(2)

Eur

p

p

p

1

v

1

where f and γp(3)

are the function of stress andplastic strains respectively. The second yield surface(cap yield surface) is taken to close the elastic region
in isotropic compression defined as follows:

f =
c

2
q

α

2

+ p′ − p p
2

2

(4)

Where α is an auxiliary model parameter that is
nc

related to K 0 .
Following [14], a principle hyperbolic function isused to express the stress-strain relations of HS
model in primary triaxial loading which can be
expressed as:

1
q
−ε1 =
for: q < q f
Ei 1 − q / qa(5)Here, ε 1 is the vertical strain and q

Fig. 1 (a) yield surface of the hardening soil modelon the p − q plane (b) stiffness dependent fromtriaxial test (c) stiffness dependent from oedometer
test

is thedeviatoric stress. qa and Ei are the asymptotic value
and the initial stiffness.
For the stiffness dependent, the primarydeviatoric loading and primary compression arecontrolled by E50 and Eoed , respectively, as shown

The plastic strain increment can be determinedfollowing relationship between εv and γ , and thepflow rule is given as:

in Figs. 1(b)-(c):εv = sin ψ m γp′ref  c cos ϕ − s 3 sin ϕ m

E50 = E50 ref c cos ϕ + p sin ϕ  c cos ϕ − s 3′ sin ϕ 
nc


K0
ref
Eoed = Eoed ref c cos ϕ + p sin ϕ 




ref

p

p

(8)

According to the original Rowe’s stress-dilatancy
theory [15], the dilatancy equation can be written as
follows:(6)m

sin ψ m =(7)

sin ϕ m − sin ϕ cv
1 − sin ϕ m sin ϕ cv

(9)where ψ m is the mobilized dilatancy angle, ϕ cv andrefwhere E50 and Eoed are the reference modulus fromϕ m are the critical state friction angle and thethe triaxial (Secant modulus) and oedometer

mobilized friction angle respectively.
199

GEOMATE- Bangkok, Nov. 14-16, 2016

σ 3 . Inspection of this figure shows that ϕ decreases

3. EXPERIMENTAL EVIDENCE AND MODEL
MODIFICATION

with an increasing of σ 3 and can be derived as:

In the previous section, the detail of HS model has
been described. To better adapt the model for
application to rockfills, based on experimental data,some additional features of which still have not yet
been considered. In this section, several key features
were identified and a new constitutive parameter isintroduced.
The experimental data (including data fromtriaxial and oedometer tests) used in this researchwere based on those for the construction of the Nam
Ngum 2 dam by the Institute of Water Resourcesand Hydropower Research (IWHR) and the Asian
Institute of Technology (AIT). The rockfills in thisproject consisted of blended rockfill materials
containing sandstone and siltstone (six types) [16] aslisted in Table 1. The confining pressures werevaried from 0.5 to 2 MPa in the triaxial tests,
whereas a maximum vertical stress of 3.2 MPa wasapplied in the oedometer tests.

 σ 3′  pa 

ϕ= ϕ 0 − ∆ϕ log 

where pa is the atmospheric pressure, ϕ 0 is thereference friction angle, and ∆ϕ is the reduction
factor.

Fig. 3 Relationship between moed and mtri

Table 1 Characteristics of rockfill materials used
Type
I
II
III
IV
V
VI

Materials
Sandstone
Sandstone
Sandstone
Sandstone: 85%,
Siltstone: 15%
Sandstone: 70%,
Siltstone: 30%
Sandstone: 55%,
Siltstone: 45%

Densities(g/cm3)

Secondly, in this work the new model parametertaking into account the stiffness dependent is derived
from the experimental data. As mentioned inprevious section, E50 and Eoed vary with the

Dmax

2.15
2.15
2.15

(mm)
40
60
60

2.15

60

2.10

60

2.10

60

(10)

evolution of the stress state in a power-law form( m) . For fine grained soils, m values wereapproximately the same values corresponding toboth triaxial and oedometer tests. The HS model
assumes the same degree of evolution with the stressstate m for both E50 and Eoed . The power
parameter m is typically determined from the resultsof triaxial tests and used for both E50 and Eoed .
However, no study has yet discussed the validity ofthese stiffness moduli for rockfill materials becausevery few authors have reported data from both
triaxial and oedometer tests. As illustrated in Fig. 3,

mtri values (from triaxial test) are higher than moedvalues (from oedometer test) the relationship of theaforementioned data can be estimated as the linear
function as:
Fig. 2 Relationship between ϕ and σ 3

moed = 0.651mtri(11)

To take an account for separate degree of stressFirst, one of the important features to beconsidered in rockfill behavior is the variation of ϕ .

dependent stiffness, the model constant moed needed
to be defined separately and denoted by n in thisstudy:

Fig. (2) shows the relationship between the ϕ and200

GEOMATE- Bangkok, Nov. 14-16, 2016

 d 

d 

3 −α c d F (d ) = d 

3 −α uF (d ) = 
c

(15)

M

(16)

u

M

where α , α c and α u are the fractal dimensions for0

the initial, current and ultimate gradation curves,
respectively, and d , d m and d M are the particle
diameter, minimum particle diameter and maximum
particle diameter, respectively. α and α c can be0

determined by fitting the gradation curve before andafter the test. For rockfill materials, a value of α =Fig. 4 Rockfill material (type I) observed at various
confining pressuresu

2.7 is generally used [19]. The particle breakage
evolution related to plastic work Wp can be definedFinally, the present paper presents a simplified
method of considering the effect of particle breakage
on the dilatancy of rockfill materials. Fig. 4 showsin terms of a hyperbolic function (see Fig. 6):

Bg =the q − ε 1 − ε v response of rockfill materials type I(RF I). Obviously, the δε / δε gradient (dilation)
v

χ +W

(17)
p

where Wp can be written in the following form:

1

decreases after the peak state with an increasing
degree of particle breakage (higher confiningpressure). The volumetric strain response of rockfill
material under shearing has been directly correlatedwith the degree of particle breakage [13], [17]. For
this present paper, a modification of Rowe’s stressdilatancy equation that incorporates particle

Wp

W
=
p

∫ p′d ε + qd ε
p

p

v

s

(18)

breakage behavior is proposed as ψ m ,bg :λsin ψ m ,bg =(sin ϕ m − sin ϕ cv ) ⋅ (1 − Bg )1 − sin ϕ m sin ϕ cv

(12)where ψ m ,bg is the modified mobilized dilatancy
angle, Bg is the particle breakage index, and λ isFig. 5 Definition of the particle breakage index [18]the power-law parameter for breakage. To quantify
the particle breakage, the modified definition
proposed by Einav (2007) [18] is adopted in thisstudy to represent the change in the particle
distribution, as shown in Fig. 5:∫ [ F (d ) − F (d )] d (log d )B =∫ [ F (d ) − F (d )] d (log d )
dM

dm

g

u

0

c

0

dM

dm

(13)

where F0 , Fc and Fu are the initial, current andultimate gradation curves, respectively: d F (d ) = 

d 
0

3 −α 0Fig. 6 Relationship between Bg and Wp

(14)

M

201

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 7 Model predictions and triaxial test results

Fig. 8 Model predictions and evolution of the particle distribution after the test
Table 2 Model parameters used for this study
Parameter

Type
III

IV

V

VI

43.23

47.09

42.60

42

43.30

2.59

1.16

2.99

2.55

2.45

3.95

I

II

ϕ 0 (degree)

46.08

∆ϕ

ψ 0 (degree)

3.2

3.2

3.2

0.5

-5

-5

ref

65

65

80

32

20

12

Eoed (MPa)

ref

52

50

55

24

17

10

m
n
Rf

0.34
0.24

0.45
0.25

0.29
0.14

0.69
0.42

0.68
0.32

0.70
0.26

0.75

0.74

0.78

0.82

0.68

0.65

Other

E

E50 (MPa)refur= 3E , p = 100 kPa, c = 1kPa, OCR = 1, K 0NC = 1 − sin ϕ , υur = 0.3
ref

ref

50

χ = 993 , λ = 0.268

Breakage parameter202

GEOMATE- Bangkok, Nov. 14-16, 2016

support of the Thailand Research fund (TRF) andKing Mongkut’s University of Technology Thonburi
through the Royal Golden Jubilee Ph.D. program,
under contract grant PHD/0006/2556. The authors
also extend their appreciation to the Office of theHigher Education Commission for granting financial
support through the Higher Education Research
Promotion and National Research University (NRU)
Project of Thailand. They are also indebted to the
CK Power Public Company Limited and Nam Ngum
2 Power Company Limited for providing thevaluable data.

4. CALIBRATIONIn this current study, the modified model was
implemented in the ABAQUS program through auser-defined material subroutine (UMAT). The 1element FEM simulation was performed in order todemonstrate the validity and the capability of themodified constitutive model. The model parameters
for NN2 rockfill materials used in FEM simulation
are listed in Table 1.
Figs. 7(a)-(c) show the comparisons between thepredicted results and test data for different rockfill
types and confining pressures. The test data arepresented by the symbols for each confining levels,
while the predicted results are displayed by the solidlines. For overall comparisons, there is a good
agreement between the predicted results and test
data. In additions, for the evolution of the particle
distribution (RF I) for various confining pressures(after test), the comparisons show that the prediction
is generally satisfactory as illustrated in Figs. 8(a)(d).

REFERENCES[1] Marsal RJ, "Large Scale Testing of Rockfill
Materials" Journal of Soil Mechanics andFoundation Engineering Division, ASCE, Vol.
93, No. SM2, 1967, pp. 27–43.
[2] Marachi ND, Chan CK, Seed HB and Duncan
JM, “Strength and deformation characteristicsof rockfill materials” Rep. No. TE 69-5, Civil
Engineering Dept., Univ. of California,
Berkeley, California, 1969.
[3] Charles JA and Watts KS, “The influence ofconfining pressure on the shear strength ofcompacted rockfill” Géotechnique, Vol. 30. No.
4, 1980, pp. 353–367.
[4] Thiers GR and Donovan TD, “Field density
gradation and triaxial testing of large-size
rockfill for little blue run dam” Laboratory
Shear Strength of Soil, ASTM, STP, 740, R. N.
Yong and F. C. Townsend, eds., ASTM,
Philadelphia, 1981, pp. 315–325.
[5] Varadarajan A, Sharma K, Venkatachalam K
and Gupta A, “Testing and modeling two
rockfill materials.” Journal of Geotechnical andGeoenvironmental Engineering, Vol. 129, No. 3,
2003, pp. 206–218.
[6] Zhang G, Zhang JM and Yu Y, “Modeling ofGravelly Soil with Multiple Lithologic
Components and its Application,” Soils andFoundations, Vol. 47, No. 4, 2007, pp. 799-810.
[7] Loupasakis CJ, Christaras BG, Dimopoulos
GCh and Hatzigogos TN, “Evaluation ofplasticity models ability to analyze typical earth
dams soil materials.” Geotechnical andGeological Engineering, Vol. 27, No. 1, 2009,
pp. 71–80.
[8] Araei A, ”Artificial Neural Networks for
Modeling Drained Monotonic Behavior ofRockfill Materials.” International Journal of
Geomechanics, Vol. 14, No. 3, 2014, 04014005.

5. CONCLUSIONSIn this study, an existing HS model is modified for
simulating the stress-strain-volumetric strain
response of rockfill material. The comprehensive
testing results of Nam Ngum 2 rockfill materials areused to investigate the significance of the effects ofthe stress-dependent stiffness and particle breakage.
Based on the previous statement and observation,according to the HS model was originally developed
for fine grained soils. For rockfill materials, themodel is developed by taking into account following
features:• The friction angle decreases following an
increasing of confining pressure.
• The degree of stress dependent stiffness
values (for triaxial and oedometer tests)
separately and defined by the newhardening parameter.
• A modification of Rowe’s stress-dilatancy
theory is proposed to account for the
influence of particle breakage inconjunction with the particle breakage
index ( Bg ) developed by Einav (2007).
The validation of the modified model was confirmedby the comparing between the predicted results andexperimental data. Overall comparisons, themodified model is able to predict reasonable well aq − ε 1 − ε v response of rockfill materials.
ACKNOWLEDGEMENTS
The authors gratefully acknowledge the financial
203

GEOMATE- Bangkok, Nov. 14-16, 2016

Mechanics and Foundations Division, ASCE,
Vol. 96, No. 5, 1970, pp. 1629–1653.
[15] Rowe PW, "The Stress-Dilatancy Relation for
Static Equilibrium of an Assembly of Particles
in Contact." Proceedings of The Royal Society A,
Vol. 269, 1962, pp. 500-527.
[16] IWHR., Report on Laboratory Tests of the
Rockfill Materials of Nam Ngum 2 CFRD.
Institute of Water Resources and Hydropower
Research, Beijing, China, 2007.
[17] Liu H and Zou D, “Associated Generalized
Plasticity Framework for Modeling Gravelly
Soils Considering Particle Breakage.” Journalof Engineering Mechanics., Vol. 139, No. 5,
2013, pp. 606–615.
[18] Einav I, "Breakage mechanics-Part II:modelling granular materials." Journal of the
Mechanics and Physics of Solids, Vol. 55, No. 6,
2007, pp. 1298-1320.
[19] Yang ZY and Juo JL, "Interpretation of sieve
analysis data using the box-counting method forgravelly cobbles." Canadian Geotechnical
Journal, Vol. 38, No. 6, 2001, pp.1201-1212.

[9] Özkuzukiran S, Özkan MY, Özyazicioğlu M
and Yildiz, GS, “Settlement behaviour of aconcrete faced rock-fill dam.” Geotechnical andGeological Engineering, Vol. 24, 2006, pp.
1665–1678.
[10] Soroush A and Araei AA, “Analysis ofbehaviour of a high rockfill dam.” Proceedingsof the ICE - Geotechnical Engineering, Vol.
159, No. 1, 2006, pp. 49-59.
[11] Mahabad NM, Imam R, Javanmardi Y and
Jalali H, “Three-dimensional analysis of aconcrete-face rockfill dam”, Proceedings of the
ICE - Geotechnical Engineering, Vol. 167, No.
4, 2014, pp. 323-343.
[12] Schanz T, Vermeer PA and Bonnier PG, “Thehardening soil model – formulation andverification.” Proceedings Plaxis Symposium"Beyond 2000 in Computational Geotechnics,
Amsterdam, Rotterdam, Balkema, 1999.
[13] Xiao Y, Liu H, Chen Y and Jiang J, “Strengthand Deformation of Rockfill Material Based onLarge-Scale Triaxial Compression Tests. I:
Influences of Density and Pressure” Journal of
GeotechnicalandGeoenvironmental
Engineering, Vol. 140, No. 12, 2014, .04014070[14] Duncan JM and Chang CY, “Nonlinear analysisof stress-strain in soils”, Journal of the Soil204

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

LONG TERM MECHANICAL AND LOAD TRANSMISSIONPROPERTIES OF SOIL-CONCRETE BLOCK
Kayo Doumoto1, Kiyonobu Kasama2, Mitsunari Hirasawa2, Kouki Zen2, Zentaro Furukawa2, Yuichi Yahiro2,
Masato Nakamichi3, Makoto Yamaguchi3, Takashi Umeyama3, Masaaki Katagiri4, Osamu Kawahara5 andToshiyuki Nagano6
1
Graduate School of Engineering, Kyushu University, Japan; 2Division of Civil and Structural Engineering,
Faculty of Engineering, Kyushu University, Japan; 3Ministry of Land, Infrastructure, Transport and Tourism,
Kitakyushu, Japan; 4Nikken Sekkei Civil Engineering Ltd., Japan; 5Coastal Development Institute ofTechnology, Japan; 6Rinkai Construction Co., Ltd., Japan

ABSTRACT
In order to recycle dredged soil, our research group developed interlocking blocks from dredged soil.
Dredged soil interlocking blocks (350 mm in width, 350 mm in depth and 100 mm-150 mm in height) weremade from soft clay dredged from the Kanmon Waterway. They were produced with the constant dewateringpressure 5 MPa and cement content of 15 %, 20 % and 25 % per the dry weight of clay. The unconfinedcompression test was carried out to investigate material characteristics of dredged soil interlocking blocks in thewater for six months and one year respectably. In order to check the effect of interlocking blocks, a road wasconstructed by using interlocking blocks on the surface. As a roadbed material, dredged and dewatered soil andcrushed stone for mechanical stabilization were used. Characteristics of dispersion of traffic load and settlement
were evaluated by running experiments. Stress was measured by four earth pressure gauges and settlement wasmeasured. The following conclusions are obtained: (1) There was close relationship between unconfinedcompressive strength and water-cement ratio. The correlation became stronger with curing period. (2) Dredged
and dewatered soil can disperse traffic load effectively. (3) When using crushed stone for mechanicalstabilization as roadbed material, soil-concrete blocks can disperse traffic load effectively after one year from theconstruction. When using dredged and dewatered soil as roadbed material, sinking suppressing effect wasobserved.
Keywords: Soil, Cement, Unconfined compressive strength, Stress, Settlement

carried out on interlocking blocks after immersed in
the water for six months and one year. A road wasconstructed by using interlocking blocks on thesurface and characteristics of dispersion of trafficload and the settlement were evaluated by running
experiments.

INTRODUCTION
The soft clay piled on the bottom of the ocean orbay is problematical against the in-service of shipnavigation. This soft clay is regularly dredged out,
but the shortage of disposal sites has become aserious problem.
At the Kanmon Waterway, an importantinternational seaway, dredging work is carried out
for deepening seabed every year. Dredged soil istransported to Shinmojioki disposal site, but thecapacity will be full in the middle of 2018 [1].
Construction of new disposal site is inevitably
necessary, but it takes time.
In order to recycle dredged soil, our researchgroup has conducted basic study and Yamashita et al.
[2] succeeded making soil block (Φ 53.4 cm × H 50cm) that has strength equals to concrete from
dredged soil.
In this study, we developed interlocking blocks
from dredged soil. Cement was mixed with soft clay
dredged from the Kanmon Waterway and the
mixture was dewatered. To investigate mechanicalproperties, the unconfined compression test was

EXPERIMENT OUTLET
Material Experiment
The material used in this study is soft clay
dredged from the Kanmon Waterway (called
Kanmon clay). The physical properties of Kanmon
clay is shown in Table 1.
Table 1 Physical properties
Kanmon clay
Soil particle density ρ s (g/cm )2.697
Liquid limit w L (%)
95.0
Plastic index I p (%)
59.1
Cement content is 15 %, 20 % and 25 % of blast
3205

GEOMATE- Bangkok, Nov. 14-16, 2016

points to investigate differences of roadbed
materials and to check the effect of interlocking

furnace slag cement type B per the dry weight ofclay. The test conditions are shown in Table 2.

Kanmon clay was prepared with initial watercontent of 300 % and mixed with cement. The
mixture was poured into the dewatering
equipment (W 350 mm × D 350 mm × H 1000mm) shown in Fig. 1.

Interlocking block

Surface
Roadbed

20cm

Base course
Earth pressure gauge

Table 2 Test conditions

Initial watercontent
Curing condition
Curing period

pistonFig. 2

A

5m

C

5m

5m

D

1.05m
5m
Empty car(10t)

Water curing (Temperature:
20℃)
6 months, 1 year

Crushed stoneA’

Dehydrated soil

B’

C’
:Track

Fig. 3

scraper launcher
Oring
polus stoneD’:Earth pressure gauge:Interlocking block

The horizontal projection of road

STRENGTH PROPERTY

Figure 4 shows the relationship between
unconfined compressive strength and cement
content. Unconfined compressive strength cured
for 6 months increased with increasing cement
content. Average unconfined compressive
strength for cement content of 20 % was 8.06
MPa (Curing period 6 months) and 10.80 MPa(Curing period 1 year). Unconfined compressive
strength increased with curing period because ofsolidification of cement.

drainage

Fig. 1

B

Loaded car(20t)drainage1374

The sectional view of road

1.05m

Kanmon clay
Blast furnace slag cement
type B15, 20, 25 %2 MPa: 30 min. → 5 MPa:
265 ~ 755 min.
300%

2.93m

Cement content
Dewatering mode

blocks. Loaded cars and empty cars were used astraffic loads. The settlement was measured.

9m

Clay
Cement type1m
(Crushed stone or Dehydrated soil)

490

1000

(mm)

The dewatering equipment

Unconfined compressive strength (MPa)The mixture was dewatered with dewateringpressure of 2 MPa for 30 min. followed byconstant dewatering pressure of 5 MPa. In order
to investigate material characteristics, theunconfined compression test (JIS A 1216) was
carried out.
Running Experiment

The sectional view of road is shown in Figure
2 and the horizontal projection of road is shown in
Figure 3.
The earth pressure gauges were set in basecourse for measuring vertical earth pressure. After
spreading each layer (20 cm), pressure was given
8 times by rolling machine (4 t.). These works
were repeated 5 times and roadbed of 1 meter inheight was constructed. Dredged and dewateredsoil (called “dewatered soil) and crushed stone formechanical stabilization (called “crushed stone”)
were used as roadbed materials. On the surface ofthe road, soil blocks were used as interlocking
blocks. The earth pressure gauges were set in four

25
Curing period 6 months
Curing period 1 year

20
15
10
5
0
10

15

20

25

30

strengthandCement content (%)

Fig. 4

206

Unconfined compressive
cement content

GEOMATE- Bangkok, Nov. 14-16, 2016

Unconfined compressive strength (MPa)Figure 5 shows the relationship between
unconfined compressive strength and dry density.
Change of dry density by passage of curing period
wasn’t observed much. Unconfined compressive
strength increased with increasing dry density
regardless of cement content. From this, increaseof unconfined compressive strength is related to
not only density but also solidification of cement.
Increase of unconfined compressive strength bycuring period was observed remarkably with highdry density for cement content of 20 %.

The
relationshipbetweenunconfinedcompressive strength and water-cement ratio isshown in Figure 7. In this paper, the water-cement
ratio is defined as weight ratio of water andcement. Unconfined compressive strength
increased with decreasing water-cement ratio.
Following equations between unconfinedcompressive strength and water-cement ratio for
cement content of 20 % were obtained:

25
Cement content 15 % (Curing period 6 months)
Cement content 20 %Cement content 25 %Cement content 20 % (Curing period 1 year)

20

qu 6month = 17.71( w / c) −0.912(1)

qu1 year = 39.22( w / c) −1.517(2)

Eq. (1) shows the result of relationship between
unconfined compressive strength and water-cement
ratio of soil blocks cured for 6 months. Eq. (2)shows the result of cured for 1 year. q u is the15

unconfined compressive strength and w/c is thewater-cement ratio. The coefficient of correlation
R in Eq. (1) and Eq. (2) were 0.69 and 0.96
respectively. Correlation became stronger withcuring period.

10
5
0
0.7

0.8

0.9

1

1.1

1.2

1.3

1.4

3

Fig. 5

Unconfined compressive strength (MPa)Dry density (g/cm )Unconfined compressive strength and dry
density

Unconfined compressive strength (MPa)Figure 6 shows the relationship between
unconfined compressive strength and watercontent. Change of water content by passing of
curing period wasn’t observed much. Unconfinedcompressive strength decreased with increasingwater content regardless of cement content. Thisis because of water left as gap. Increase of
unconfined compressive strength by passing of
curing period was observed remarkably with low
water content for cement content of 20 %.

25
20
15
10

Fig. 7
Cement content 15 % (Curing period 6 months)
Cement content 20 %Cement content 25 %Cement content 20 % (Curing period 1 year)

5

50

60

70

80

90

100

Water content (%)

Fig. 6

1

2

3

4

5

6

7

Unconfined compressive
water-cement ratio

strengthandFigure 8 shows the relationship between heightof roadbed and vertical stress measured in basecourse when constructing roadbed. As the height ofroadbed grew, vertical pressure increased linearly.
Increasing rates of stress were 27.96 kPa/m (crushedstone) and 3.91 kPa/m (dewatered soil). The
increasing rate of vertical stress for crushed stone
was 7 times higher than that of dewatered soil. The
increasing rates of stress calculated by wet unit
weight were 17.36 kPa/m (crushed stone) and 11.23
kPa/m (dewatered soil). Therefore, it is considered

10

40

0

LOAD TRANSMISSION

15

0
30

5

Water-cement ratio

25
20

Cement content 15 % (Curing period 6 months)
Cement content 20 %Cement content 25 %Cement content 20 % (Curing period 1 year)

Unconfined compressive strength and watercontent
207

GEOMATE- Bangkok, Nov. 14-16, 201635

Vertical stress (kPa)30

that dewatered soil can disperse traffic load
effectively.
Fig. 8Height of roadbed and vertical stress

The relationship between vertical stress andrunning frequency right after the construction andafter one year from the construction are shown in
Figure 9 and Figure 10 respectively. From Fig. 9,vertical stress was about 30 kPa smaller when
dewatered soil was used as roadbed compare to thecase when crushed stone was used. And the verticalstress became larger when soil blocks were set on
the surface of the road. Vertical stress where blocks
were set on the surface was 1.04 times (crushedstone) and 2.83 times (dewatered soil) larger
compare to where no block was set on the surface[3]. From Fig. 10, vertical stress became smaller0

Height of roadbed (cm)27.96kPa/m

5
10
15
20
Vertical stress (kPa)

25

30

when using soil blocks on the surface and crushed
35

Vertical stress (kPa)

30
25
20

Crushed stone + Block

Crushed stone

15
10

Dewatered soil + Block

Dewatered soil

0
2

4
6
8
10
12
Running frequency (times)

10
20
30
40
Running frequency (times)

50

14

Traffic stress amplitude (kPa)0stone was used as roadbed. Vertical stress where
blocks were set on the surface was 0.85 times(crushed stone) and 1.97 times (dewatered soil)
larger compare to where with no block setting on thesurface. From Fig. 9 and 10, change of vertical stresswasn’t observed much when using dewatered soil as
roadbed and negative pressure wasn’t measured one
year from the construction. This is because ofstability of the foundation.
Fig. 9050

Fig. 11 Traffic stress amplitude and runningfrequency (right after the construction)

5-5

Dewatered soil

The relationship between traffic stress amplitude
and running frequency right after the construction
and after one year from the construction are shownin Figure 11 and Figure 12 respectively. Traffic
stress amplitude is defined as the gap of maximum
stress and minimum stress measured by traffic load.
Fig. 11 shows that traffic stress amplitude increases
where with block setting on the surface. Traffic
stress amplitude tended to decrease due to increaseof traffic frequency. The traffic stress amplitude wassmaller when dewatered soil was used as roadbed
compare to the case when crushed stone was used[3]. From Fig. 12, traffic stress amplitude also
increased when setting blocks on the surface. Traffic
stress amplitude where with blocks on the surface
was 2.32 times (crushed stone) and 3.09 times(dewatered soil) larger compare to when setting noblock on the surface. Traffic stress amplitude tended
to converge constant value compare to Fig. 11 anddeference of whether soil blocks were set or not
could observe clearly.

100-5Dewatered soil + Block

5

Fig. 10 Vertical stress and running frequency (after
one year from the construction)

80

120

10

after the construction)

40
3.91kPa/m

15-5

Crushed stone
Dewatered soil

60

Crushed stone

20

0

0
20

Crushed stone + Block

25

208

Dewatered soil + Block

40
30
20
Crushed stone

Dewatered soil

100Running frequency and vertical stress (rightCrushed stone + Block

0

10
20
30
40
Running frequency (times)

50

GEOMATE- Bangkok, Nov. 14-16, 2016

CONCLUSION

Traffic stress amplitude (kPa)50
Dewatered soil + Block

In order to investigate mechanical and load
properties of soil interlocking blocks, the unconfinedcompressive strength was examined and running
experiments were carried out. The following
conclusions are obtained: 1) There was close
relationship between unconfined compressive
strength and water-cement ratio. The correlation
became stronger by passage of curing period. 2)
Dredged and dewatered soil can disperse traffic load
effectively. 3) When using crushed stone formechanical stabilization as roadbed material, soilconcrete blocks can disperse traffic load effectively
after one year from the construction. When using
dredged and dewatered soil as roadbed material,
sinking suppressing effect was observed.

40
30
Crushed stone + Block20
10

Dewatered soil

0

0

2

Crushed stone

4
6
8
10
12
Running frequency (times)

14

Fig. 12 Traffic stress amplitude and runningfrequency (after one year from theconstruction)

ACKNOWLEDGEMENTS

The relationship between settlement and runningfrequency just after the construction is shown in
Figure 13. The settlement increased with theincrease of running frequency. Especially, thesettlement was large when dewatered soil was usedas roadbed. The effect of soil block could not be
confirmed when using crushed stone as roadbed
material, but when using dewatered soil, soil block
reduced the settlement more than 1 cm. This isbecause of occlusion of blocks.

We appreciate the generous support received
from SCOPE for research development of this study.
We gratefully acknowledge the “Shin-Mojioki
disposal site construction project III technology
investigative commission” for their professional
advice. We also express our deepest gratitude to all
who were involved in this study for their time andhelp.
REFERENCES[1] Ministry of Land, Infrastructure, Transport and0

Crushed stone

Tourism Kyushu Regional Development Bureau,“Shin-Mojioki Disposal Site (Second Period)
Public Ownership Water Surface Land Fill
Project Environmental Impact Assessment
Preliminary Documents”, 2015, 2-1.

Settlement (cm)0.5
Crushed stone + Block1
1.5[2] Yamashita Y, Kasama K, Zen K and Chen G,

2

Dewatered soil + Block“Dehydration and Strength Properties of
Cement-mixed and Dehydrated Clay”, The
International Conference on Advances inGeotechnical Engineering (ICAGE 2011), 2011,
pp. 151–155.

2.5
3
3.5

Dewatered soil

0

10
20
30
40
Running frequency (times)

50[3] Doumoto K, Kasama K, Hirasawa M, Zen K,
Furukawa Z, Yahiro Y, Nakamichi M,
Sakamoto M, Ueda T, Katagiri M, Yamamoto S,
Kawahara O, and Nagano T, “Mechanical andLoad Transmission Properties of Soil-concrete
Block”, 6th Japan-China GeotechnicalSymposium, 2015.

Fig. 13 Settlement and running frequency (just after
the construction)209

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EFFECTS OF VERTICAL WALL BARRIER DUE TO RIGIDPAVEMENT DEFLECTION OF FULL SCALE 1-PILE ROWNAILED-SLAB SYSTEM ON SOFT SUB GRADE
BY COMPRESSION LOADINGS
Anas Puri1,a, Hary C. Hardiyatmo2,b, Bambang Suhendro2,c, Ahmad Rifa’i3,d1

Associate Professor Department of Civil Engineering, Islamic University of Riau, Indonesia
Professor Department of Civil and Environmental Engineering, Gadjah Mada University, Indonesia
3
Associate Professor Department of Civil and Environmental Engineering, Gadjah Mada Universityaanaspuri@eng.uir.ac.id, bharychistady@yahoo.com, cbbsuhendro@yahoo.com, dahmad.rifai@ugm.ac.id2

ABSTRACTThe weaker part of rigid pavement when it is loaded is the edge of slab. Vehicle wheels that are often runningin/out from/to pavement can cause void in the sub base-pavement interface for a long time loading period.
Flexural moment caused by temperatures also precipitate the damages of pavement edge. To avoid the damageof pavement edge, the vertical wall barrier is added. This research is aimed to learn the contribution of verticalwall barrier on the pavement of Nailed-slab System to reduce the slab deflection. Full scale observation was
done over models in soft clay soil. The full scale of 1 pile row Nailed-slab System was conducted on soft clay
which consisted of 6.00 m x 1.20 m slab area with 0.15 m in slab thickness, 5 short micro piles (0.20 m indiameter, 1.50 m in length, and 1.20 m in pile spacing) as slab stiffeners which were installed under the slab.
Piles and slab were connected monolithically, then in due with vertical concrete wall barrier on the two ends ofslab. The system was loaded by compressive loadings on the center and the end of slab. Deflection of model
without vertical wall barrier was analyzed by finite element method. Results show that the vertical wall barriercannot significantly reduce the deflection for edge loading. It is opposite to the model test result from Puri, et.al.
[1] where the vertical wall barrier can reduce 74% deflection for edge loadings. It is to be expected thatnumerical application program could not model the vertical wall barrier which lower position than slab level.
Keywords: rigid pavement, soft clay, Nailed-slab System, vertical wall barrier, deflection.

INTRODUCTION

grade and pavement due to the effect of the tires
which often exit/ enter to/ from pavement, (3) toreduce the disturbing on berm, and (4) as a verticalwall barrier to isolate the negative effects of waterchanging in sub grade. It prevents the water
infiltration to the soil under the pavement slab.

The edge of a rigid pavement is the weaker partin bearing the traffic load. This part tends to get
maximum deflection and to damage the pavement
such as broken pavement, developed voids between
sub grade and pavement, and can be followed bypumping. The edge of rigid pavement is also theweaker part in the Nailed-slab System [1]. Hence, inthis research the edge of the pavement is reinforced
by a vertical wall barrier. Furthermore, the behaviorof Nailed-slab System and the contributions ofvertical wall barrier will be studied.
A vertical wall barrier is usually reinforced
concrete that is 10 cm – 20 cm in thickness and it isconducted in a vertical position that is 40 cm – 50
cm in height. Figure 1 shows an example of avertical wall barrier. Some benefits of a vertical wallbarrier are [2]: (1) it acts as a slab stiffener for edgeof pavement because this part will be a weaker partof pavement, (2) to avoid the void between sub

Fig. 1 An example of vertical wall barrier [3] (TheIndonesian Research and Development Center for
Highway and Bridge, 2007)
Puri, et.al. [1] (2011) conducted the 1-pile row
Nailed-slab model tests on the soft clay to learn theeffects of vertical wall barrier. The vertical wall

210

GEOMATE- Bangkok, Nov. 14-16, 2016

Then the instrumentations were recorded. Details
about testing procedure is presented in Puri, et.al.
([4], [5]). Some photographs in construction andtesting are presented in Fig. 2.

barrier can reduce deflection 34% and 74% forconcentric and edge loadings respectively.
This paper is aimed to discuss the comparison of
the slab deflection behavior between the Nailed-slab
with a vertical wall barrier and without a verticalwall barrier. The Nailed-slab models consider thevertical wall barrier on each end of the slab. The
experimental study was conducted by full scalemodel tests of one row pile Nailed-slab system andanalyzed by finite element method.

Table 1 Soft Clay Properties [4]Parameter
Spesific gravity, G s

Unit

Average-

2.55

Consistency limits:
INVESTIGATED 1-PILE ROW FULL SCALE
NAILED-SLAB- Liquid limit, LL

%

88.46

- Plastic limit, PL

%

28.48

Detail of the procedure on 1-pile row full scale
Nailed-slab is presented in Puri, et.al. [4] and briefly
described in Puri, et.al. [5]. In this paper, it will bepresented again comprehensively.

- Shrinkage limit, SL

%

9.34

- Plasticity index, PI

%

59.98

- Liquidity index, LI

%

0.36

Natural water content, w n

%

50.49

Soil Pond and Materials

Water content, w

%

54.87

A full scale model of Nailed-slab was conductedon soft clay. A 6 m x 3.6 m soil pond was conductedby digging the existing soil until the depth of 2.5 m.
On the two longer sides it was retained by masonrywalls and supported by some temporarily bamboo
girder. The anchorage system was built near the
pond. Separator sheets were set on the pond walls
and base to avoid the effects of surrounding existingsoils. A 2.15 m of pond depth was filled by soft clay
which was taken from District Ngawi, East Java,
Indonesia. The soft clay properties are presented inTable 1. The slab and piles were reinforced concrete.
The concrete strength characteristic of the slab andpiles was 29.2 MPa and 17.4 MPa respectively. The
flexural strength of the slab was 4,397.6 kPa.

Clay content
Sand content
Bulk density, γ
Dry density, γ dUndrained shear strength, s u- Undisturbed- Remolded
CBR
Soil classification:- AASHTO- USCS

%
%
kN/m3
kN/m3

92.93
6.89
16.32
10.90

kN/m2
kN/m2
%

20.14
11.74
0.83

-

A-7-6
CH

Dimension of Nailed-slab Prototype

In the 2D finite element analysis (FEM), MohrCoulomb soil model was employed in the study.
Likewise, soil parameters and idealization of
structural elements are presented in Table 2 and 3,
respectively. The material properties were adjusted
due to the plain strain case [6]. The slab width is 120
cm and the length of considered section is 20 cm(perpendicular to cross section). Numerical analysis
was conducted by 2D Plaxis version 8.6. The soft
clay was modeled by Mohr-Coulomb in undrained
condition. All structural elements were modeled byplate element in linear-elastic behavior. Lean
concrete was modeled by soil with linear-elastic
non-porous material. The thickening slab was
ignored since it could not be modeled by numericalapplication program.
The used mesh in plain strain FEM analysis isshown in Fig. 3. Fig. 4 shows one of deformed
shape outputs.

Analysis of Deflections

The Nailed-slab System Prototype dimension was
6.00 m × 3.54 m, 0.15 m in slab thickness, and theslab was reinforced by micro piles 0.20 m indiameter and 1.50 m in length. The spacing between
piles was 1.20 m. This model was obtained by
cutting the 600 cm × 354 cm × 15 cm Nailed-slab to
3 parts where each part consisted of one pile row.
The tested 1 pile row Nailed-slab was the middle
one with slab dimension 600 cm × 120 cm × 15 cm
as shown in Figure 2. All piles were installed under
the slab and connected monolithically by using
thickening slab connectors (0.40 m × 0.40 m and0.20 m in thickness). Each end of slab is equippedby the vertical concrete wall barrier. There was a 5
cm lean concrete thickness under the slab. The slab
was loaded by compression loadings with different
load positions. Loads were transfered to the slab
surface by using a circular plate 30 cm in diameter(the plate represents the wheel load contact area).

211

GEOMATE- Bangkok, Nov. 14-16, 2016

AnchorAnchor beam

Masonry wall

Soft clay

104

354

60
60
Concrete pile

50

60

107

120

120

120

120
Slab, h=15

60 50a). Plan of view
Loading beam
Instrumentation:
Pressure meter
Dialgauge

Anchor beam

Steel clamp
Hydraulic jackon loadcell

Lean concrete, 5 cm

II II

I

Reference beam

Vertical barrier,d = 12.5
50
15
15
20

50

150

Soft clay

50

Pond base-medium sand50
Concrete pile,
Thickening slab,
D= 20, L = 150
40×40, h= 20

600

50

Concreteslab, h=15

Unit in cm. Without scaleb). Cross section

Fig. 2 Schematic diagram of full scale Nailed-slab with 1 pile row.

Table 2 Model and parameters of soilParameters
Material model
Material behavior
Saturated density
Dry density
Young's Modulus
Poisson's ratio
Undrained cohesion
Internal friction angle
Dilatancy angle
Initial void ratio
Interface strength ratio

Name/ Notation
Model
Type

γ sat
γd
E
v
cu

φ
ψ

e0
R

Soft clay
Mohr-Coulomb
Undrained
16.30
10.90
1,790.00
0.45
20.00
1.00
0.00
1.19
0.80
212

Sand
Mohr-Coulomb
Drained
18.00
20.00
42,750.00
0.35
1.00
47.80
2.00
0.50
0.70

Unit
kN/m3
kN/m3
kPa
kPa
o
o
-

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 3 Model and parameters of structural elements in FEM 2D plain strain
Parameters

Name/
Notation

Lean concrete(LC)

Material model

Model

Material behavior
Normal stiffness
Flexural rigidity
Equivalent thickness
WeightPoisson's ratio
DensityYoung's ModulusInterface strength ratio

Type
EA
EI
d
w
v

Volumeelement
Elastic
-

γ

E
R

0.2
22
17,9000.80

Structural elements
Slab

Unit

Plate

Verticalwall barrier
Plate

Pile
Plate

-

Elastic
4,554,000
8,539
0.15
3.60
0.15
24
25,3000.80

Elastic
3,795,000
4.941
0.125
3.00
0.15
24
25,3000.80

Elastic
616,696
75,655
0,027
29.12
0.20
24
19,6000.80

kN/m
kNm2/m
m
kNm/m
kN/m3
MN/m2
-

uplifted for concentric loading (Fig. 5a). However,
the opposite end of the slab was uplifted caused byedge loading (Fig. 5b).

Deflection (mm)Distance from load (m)Fig. 3 A used mesh in plain strain FEM analysis.

-3

-2

-4
-1 -2 0
0
2
4
6
8

1

2

3

FEM 2D
Observeda)Deflection (mm)

-4
-2 0
0
2
4
6
8

Fig. 4 A deformed shape output of numericalanalysis for edge loading.

Distance from load (m)

2

4

6

FEM 2D
Observedb)
Fig. 5 Deflection shape along the slab for load P =80 kN; a) centric load, b) edge load.

RESULTS AND DISCUSSION
P-δ Relationship for Nailed-Slab with Vertical
Wall Barrier

According to Figure 6, the maximum shear
between soil and pile occured around the pile tip ofthe middle piles (Fig. 6a) and the left pile for edge
loading (Fig. 6b). The extreme shear stresses on pile
were 12.63 k/m2 and 10.46 kN/m2 for concentricloading and edge loading, respectively. Both shear
stresses did not reach the ultimate value of 16.00
kN/m2 (=R inter c u ).

Fig.5 shows deflection shape along the slab
(cross section in field condition). It is seen that the
deflection shape is a bowl shape. Analysis results
show that maximum displacement occured under theloading point and decreased as you move further
away from the loading point. This phenomenon
fulfilled the expectation. The maximum deflections
were appropriate with the observations. But the endof slab was uplifted. Both ends of the slab were213

GEOMATE- Bangkok, Nov. 14-16, 2016

is opposite to the model test result from Puri, et.al.
[1] where the vertical wall barrier can reduce 74%
deflection for edge loadings. Soil shear stresses did
not reach the ultimate value for both load positions.

8

ACKNOWLEDGEMENTS

0a
2

The authors acknowledge the financial support
from the Directorate General of Research andTechnology and Higher Education Republic ofIndonesia number SP DIPA-023.04.1.673453/2015.

0

REFERENCESb[1] Puri, A., Hardiyatmo, H.C., Suhendro, B., andRifa’i, A., 2011, Contribution of Wall Barrier toReduce the Deflection of Nailed-Slab System inSoft Clay, Proc. of 9th Indonesian Geotech.
Conf. and 15th Annual Scientific Meeting(KOGEI IX & PIT XV) HATTI, HATTI, Jakarta,
7-8 December 2011, pp. 299-306 (in Bahasa
Indonesia).
[2] Suhendro, B., dan Hardiyatmo, H.C., 2010,
Modified Chicken Foot Pavement System as an
Alternative Solution for Road Construction on
Soft Soils, Expansive, and Embankment, Proc.
of One Day Seminar and Exhibition 2010 New
Innovative Technology on Highway and Bridge,
DPD HPJI Jatim, Surabaya, 31 March 2010 (inBahasa Indonesia).
[3] The Indonesian Research and Development
Center for Highway and Bridge, 2007, Uji Coba
Skala Penuh Teknologi Cakar Ayam Modifikasi,
Bab-7 Gambar Rencana, Satuan Kerja Pusat
Litbang Jalan dan Jembatan, Departemen
Pekerjaan Umum Republik Indonesia.
[4] Puri, A., Hardiyatmo, H.C., Suhendro, B., andRifa’i, A., 2013, Behaviour of Full scale Nailedslab System with Variation on Load Positions,
Proc. 1st Intl. Conf. on Infrastructures
Development (ICID), Surakarta, 1-3 November
2013, pp. 26-36.
[5] Puri, A., Hardiyatmo, H.C., Suhendro, B., andRifa’i, A., 2014, Behavior of Nailed-slab
System on Soft Clay Due to Repetitive
Loadings by Conducting Full Scale Test,
IJCEE-IJENS, Vol. 14 No. 06, pp. 24-30.
[6] Plaxis 3D Foundation, Material Models Manual
version 1.5.

Fig. 6 Shear stress on the interface of structuralelements; a) concentric load, P = 150 kN, b) edge
load, P = 80 kN.
Effects of Vertical Wall Barrier
Figure 7 shows the P-δ relationship for edge
loading. It seems that the vertical wall barrier can
reduce deflection 5.25% only. It is an insignificant
reduction if it is compared to the model test. The
vertical wall barrier can reduce 74% deflection foredge loadings in model test [1].
0Deflection (mm)02Load, P (kN)

50

100With verticalwall barrier
W/o verticalwall barrier
Observed

4
6

8
Fig. 7 P-δ relationships for load on C (edge load).

CONCLUSION
In this paper, observations for pavement of
Nailed-slab System which are equiped by verticalwall barrier on the end of slab were conducted andnumerical analysis for both systems with andwithout vertical wall barrier has been done.
Numerical results show that the vertical wall barriergave insignificant reduction to the slab deflection. It

214Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

AN EVALUATION OF OSMOTIC TECHNIQUE UNDERULTRAVIOLET GERMICIDAL IRRADIATION EXPOSURE
Mohd Yuhyi Mohd Tadza1, Nurhidayah Mahazam1 and Snehasis Tripathy2
Faculty of Civil Engineering and Earth Resources, Universiti Malaysia Pahang, Malaysia1 School ofEngineering, Cardiff University, United Kingdom2

1

ABSTRACTThe osmotic technique, which uses polyethylene glycol (PEG) solutions of varying concentrations withsemipermeable membranes of different molecular weight cutoffs (MWCO), is commonly used to apply suction
in soils. Cellulose acetate membranes which are most commonly used, are susceptible to microbial attacks. Thisin turn will lead to the intrusion of PEG into soil specimens. Osmotic and vapour equilibrium techniques areoften used to establish drying suction-water content soil-water characteristic curves (SWCC). In this study,
suctions of 0.11 to 300 MPa were applied on Andrassy bentonite slurries. At higher applied suctions, the osmotic
tests were carried under short length ultraviolet germicidal irradiation (UVGI). In addition, Atomic ForceMicroscopy (AFM) and Fourier Infrared (FTiR) were employed to evaluate the changes in the semipermeable
membranes and PEG molecules, respectively. The water content of the clay obtained from the osmotic tests wasfound to be greater at the overlapping suction region. Interestingly, under UVGI exposure, the water content was
found to be in good agreement with the water content determined using the osmotic technique at low suctions
and the vapour equilibrium technique at higher suctions. FTiR spectrum and AFM results revealed that somechanges had occurred on both the PEG and in the membrane pore sizes. However, these changes did not affectthe final water content in the bentonite and therefore, more precise suction-water content SWCC for the clay
could be established.
Keywords: Clay, Suction, Osmotic, Microbes, PEG, UV, Cellulose Acetate

osmotic techniques have gained widespread
acceptance as reliable methods for controlling
suction in soil specimens [11]. The vapour
equilibrium
technique,which
utilizes
thermodynamic principles to control the relative
humidity of various acid and salt solution vapours ina closed system, is used to apply suction in soil [9].
In the osmotic technique, on the other hand, the soil
specimen is placed in semipermeable membrane andis brought into contact with polyethylene glycol(PEG) of predetermined concentration. In thistechnique, two polymers are commonly used,
namely aqueous solution of (PEG) and cellulose
acetate semipermeable membrane [12].

1.0 INTRODUCTION
Bentonite is expansively used in various geoenvironmental applications. Bentonite in various
forms such as loose powder, compacted and slurry
have been used in the construction of liners,
engineered barriers, isolation walls and backfill
material [1][2]. Bentonites have been considered dueto their unique containment properties (i.e. low
permeability, high sorption capacity, self-sealing
membrane characteristics, and durability in a naturalenvironment) [3][4]. As of late, the study of soil
microbes and their interactions in bentonite buffer
has been of interest in nuclear waste repository
applications [5]. Previous studies have shown thatthe presence of microbes can affect the geochemistryof bentonites and thus, influence their properties andengineering behavior [6][7][8][5].

One of the main drawbacks of the osmotic
technique when using PEG 6000 along withmolecular
weight cut-off (MWCO) 3500
semipermeable membrane, is the intrusion of PEG
solution into the soil specimen particularly at higher
applied suction [13]. This problem occurs due to thebreakage of the semipermeable membrane by soil
fungi [14]. As soil hosts a diversity of microbes, thebreakage of the semipermeable membrane inosmotic tests is found to be inevitable. Removal ofsoil microbes prior to conducting osmotic tests istherefore crucial to ensure precise determination ofsoil-water SWCCs.

The engineering behavior of unsaturated soils(viz. shear strength, volume change, permeability)
due to changes in the water content are commonly
predicted by establishing the suction-water content
soil-water characteristic curves (SWCCs). In thelaboratory, the suction-water content SWCCs arecommonly established using various techniques topredict the changes in the water content in soils[9][10]. Recently, the vapour equilibrium and215

GEOMATE- Bangkok, Nov. 14-16, 2016

The main focus of this paper is to evaluate theuse of short-length ultraviolet germicidal irradiation(UVGI) to remove soil microbes prior to conducting
osmotic tests.
In this study, the suction-watercontent SWCC for Andrassy bentonite wasestablished using both osmotic and vapour
equilibrium
techniques.
The
microbial
characteristics of the bentonite were established
before the removal of said microbes by exposing
both PEG solutions and the soil specimens to UVGI.
Atomic Force Microscopy (AFM) and Fourier
Transform Infra-red (FTiR) spectroscopy were
employed before and after exposure of UVGI. todetermine if there any changes to PEG moleculesand cellulose semipermeable membrane as a resultof the irradiation.

The magnitude of the alteration was found to besignificant for lower MWCO membranes.
2.2 Intrusion of PEG
Soil microbes (i.e. bacteria and fungi) exist in
both naturally occurring and commercially available
bentonites [8][5]. The biodegradation of cellulose
acetate based material by soil microbes has beenstudied for a long time by number of authors[16][17]. It is anticipated that, these microbes were
responsible for the degradation of the membrane.
The antibacterial property of PEG was found to be
insufficient in removing soil microbes [14]. In
geotechnical application, Kassif & Ben Shalom [18],recommended the usage of penicillin duringlaboratory tests to eliminate the bacteria to ensure
long-term performance of cellulose acetate
membranes. However, a recent study revealed thatthe breakage of the membrane is due tobiodregadation of cellulose acetate by soil fungi[14]. Mohd Tadza et al. [14] also noted that the useof penicillin is ineffective in removing soil fungi.
Thus, other methods of disinfection may be required
to eliminate the presence of cellulose degrading
fungi in soil specimens.

2.0 LITERATURE REVIEW
2.1 Suction-Water Content SWCC
PEG is a hydrophilic polymer made up ofmolecular chains having the chemical formula: HO[CH2-CH2-O]n-H [9]. The use of PEG–water
mixtures along with suitable semipermeable
membranes enables the application of osmotic
gradients in an osmotic system. During thecommencement of an osmotic test, a soil specimen is
enveloped by a semipermeable membrane andimmersed in an aqueous solution of PEG of varying
concentrations. Cellulose acetate, an acetate ester ofcellulose (C6H7O2(OH)3) is commonly used for thispurpose [14]. In some cases, polysulfonate
membrane is also used [9]. The semipermeable
membrane is expected to prevent the passage of
PEG molecule to soil–water systems. A review ofliterature suggested that, the osmotic technique has
been successfully used for applying suctions in soils
up to 1.5 MPa. Although the technique can befurther extended to 12 MPa using smaller molecularweight PEGs (i.e. PEG 1500) [12], the application ofosmotic technique at higher applied suction appears
to be limited to PEG 6000 only. This could be due tounavailability of lower MWCO semipermeable
membranes [14]. For suctions higher than 10 MPa,
the vapour equilibrium technique is often opted for.
The vapour equilibrium technique has been used toapply total suction from 3 up to 1000 MPa [9][11].

2.3 Ultra Violet Germicidal Irradiation
The application of short-length UVGI in
disinfection and removal of microbes is widely used
and accepted in the microbiology community [19].
This method was also applied in disinfection of
medical instruments and operation theaters [20]. The
UV exposure at 200-280 nm for a certain period oftime damages cell’s DNA and RNA, thus destroying
the ability of the microbes to reproduce [21]. The
germicidal effectiveness peaks at UVC, which is thesubdivision of the UV wavelength (between 260-265nm) [22]. Various types of soil bacteria and fungi
were completely eliminated when exposed directly
to UV lamp under laboratory conditions [23][21].
The main concern of using UVGI in osmotic
technique is that the UV exposure may also lead todegradation of both the PEG molecules and cellulose
acetate membrane. PEG can be photochemicallydegraded when exposed to UV. However, prolonged
exposure is required. Das & Gupta (2005) noted thatsome changes in the FTiR spectrum of PEG when
exposed under UV lamp for more than 1 hour. In
the presence of oxidation chemicals such ashydrogen peroxide and titanium oxide, thedegradation of PEG in aqueous form was found tooccur at shorter periods of exposure [25][26].

In order to obtain a continuous suction-watercontent SWCC that covers a wide range of suction
for any given soil, suction-water content from bothtechniques are commonly plotted together. Studies
in the past have shown disagreement between thetest results from both vapour equilibrium andosmotic tests at higher applied suctions [15][10].
The differences were mainly attributed to alterationsof the membrane pore sizes which enabled thecrossing of PEG molecules into soil specimens [13].

Direct degradation of cellulose acetate underUVGI has not been reported elsewhere. However,previous researches have shown that, cellulose
acetate based materials were found to degrade under216

GEOMATE- Bangkok, Nov. 14-16, 2016

UV exposure at different wavelengths (i.e. 275-390nm) in the presence of photo sensitizers andoxidation catalysts [17]. Thus, the objective of thisstudy is to explore the effect of UVGI exposure
within this context.

PEG 6000 and Spectra/Por membranes with MWCO
value of 3500 were used for applying higher
suctions. Specimens for the osmotic tests wereprepared using the bentonite slurry in cellulotic
tubes of the semipermeable membranes (MWCO 14000 and MWCO 3500). Both ends of thesemipermeable membrane tubes were then securely
fastened using polypropylene clips prior to
immersing the specimens in PEG solutions. For
applied suction higher than 3.0 MPa using PEG
6000, duplicate tests were conducted under UVGI
exposure. For this purpose, the tests were conductedusing Esco Airstream Horizontal Laminar Flow
Clean Bench (Glass Side Wall). The UV lamp
automatically switched off after 10 minutes. For
establishing suction-water content relationship using
vapor equilibrium, bentonite slurries were initiallyprepared in 38 mm diameter stainless steel rings.
Once prepared, the slurry specimens were carefully
extruded and transferred directly into test desiccators.

3.0 EXPERIMENTAL TECHNIQUES
3.1 Determination of Geotechnical andMicrobiological Properties of Bentonite Samples
The physical and microbiological properties of
the Andrassy bentonite was first determinedfollowing standard laboratory procedures describedin [10][14]. The microbiological properties of thebentonite, namely bacteria and fungi determination
were carried out following plating, slide culture,streaking and isolation techniques [19]. Potato
dextrose agar (PDA) was used for culturing fungi,
whereas Nutrient agar (NA) was used to culture
bacteria. The clay specimen was initially suspended
in 0.9% NaCl solution to separate the microbes fromthe soil [27]. Isolation of each microbe was carriedout in an independent laboratory using thepolymerase chain reaction (PCR) protocol andidentification of the specific strain was done byreferring to the international microbiologicalcharacterization database.

3.3. AFM Analyses
After completion of the osmotic tests, themembranes were carefully removed and rinsed
before being directly transferred to a Nanowizard II
scanning probe AFM from JPK Instruments formorphological scanning.

3.2
Establishing the drying suction-watercontent SWCC

3.4 FTiR Analyses
A PerkinElmer Spectrum 100 spectrometer wasused for studying the chemical and molecularchanges of the PEG solutions. FTIR spectroscopy
generates spectrum patterns according to the infraredintensity (i.e. wavenumber or wavelength) and thefraction of infrared transmitted by certain chemicalbonds [28]. The spectrum patterns were theninterpreted following methods described byMorrison & Boyd [29]. A sudden shift or distortion
to the spectrum will generally indicate degradationor changes to either chemical bonds or molecule
structure [13]. PEG 6000 solutions at the end ofosmotic tests (i.e. with and without UVGI exposure)
were collected and tested.

Bentonite slurry was initially prepared by mixing
bentonite powder thoroughly with deionized water
to slightly greater than liquid limit (i.e. 1.2 x LL).
The bentonite–water mixture was stored in sealed
plastic bags and kept in an air-tight container toallow for water equilibration to take place for about
7 days prior to preparing the specimens for thelaboratory tests.
Osmotic tests were carried out by applying
suctions of 0.15, 0.25, 0.42, 1.03, 1.7, 3.65, 5.08,
6.64, 7.74, and 9.96 MPa using two types of PEG.
PEG 20 000 was used for applying lower suctions up
to 2.67 MPa, whereas PEG 6000 was used for
applying suctions greater than 2.67 MPa. The
applied suctions in vapor equilibrium were 3.6,
14.58, 23.58, 39.38, 111.77 and 262.75 MPa
corresponding to the relative humidities (RHs) of
saturated solutions of K2SO4, KNO3, KCl, NaCl,
K2CO3, and LiCl, respectively. The suctions of each
PEG and saturated salt solutions were measured
using a chilled mirror dew-point hygrometer.
Decagon WP4C was used for this purpose.

4.0

RESULTS AND DISCUSSION

4.1 Geotechnical and Microbiological Properties
of Andrassy Bentonite
The properties of Andrassy bentonite is presented inTable 1. The bentonite in this study was found to
exhibit a large surface area and high surface charge
characteristics which make it ideal for soil microbes[34]. Referring to Table 1, it was found out thateight microbes were present within the soil, (i.e.
Bacteria: Bacillus anthracist, Staphylococcus aureus,

PEG 20 000 and Spectra/Por semipermeable
membranes with MWCO value of 14 000 were used
for applying suctions lower than 3.0 MPa, whereas
217

GEOMATE- Bangkok, Nov. 14-16, 2016

membrane pore sizes.

Micrococcus luteus, Achromobacter Xylosoxidans;
Fungi: Paecilomyces lilacinus, Trichoderma
atroviridae, Fusarium proliferatum, Rhodotorulamucilaginosa). All eight microbes are considered tobe common soil microbes. Interestingly, two strainsof fungi (i.e. Paecilomyces lilacinus andTrichoderma atroviridae) found in the soil specimen
has the potential to degrade cellulose [14].
Remarkably, no presence of microbes were observed
on both PDA and NA plates after UVGI exposure,indicating that complete removal of microbes wereobtained after 10 minutes of UVGI exposure.

The drying suction-water content SWCC isshown in Fig. 2. The final water contents of the soil
specimens were found to decrease with an increasein the applied suctions. Smooth curve joining theosmotic test at lower suctions (i.e. PEG 20 000), andhigher suctions (i.e. vapor equilibrium). However,some disagreements were noted between osmotic
test results (see PEG 6000 without UVGI exposure)and vapor equilibrium test results. The osmotic tests
results at the overlapping suction region were found
to remain lower than that of water contents obtained
from vapour equilibrium. Similar observation wasreported by Tripathy et al. [10] for other types ofbentonites. The lower water contents are primarily
attributed to the intrusion of PEG solution into thebentonites specimens. On the contrary, watercontents obtained using PEG 6000 that has
undergone UVGI were found to complement the test
results of both osmotic test at lower suctions andvapour equilibrium at higher applied suctions.

Table 1 Geotechnical properties of Andrassy
bentonite
Geotechnical properties
Specific gravity, Gs
Liquid limit, wl (%)
Plastic limit, wp (%)
Shrinkage limit, ws (%)

2.78
129.3046.12
34.00

Specific surface area, S (m2/g)
Cation exchange capacity, B(meq/100g)

734.27
42.77

4.2 Suction-water content SWCC
Figure 1 shows the equilibration plot at applied
suction of 3.65 MPa using PEG 6000. Some
differences were noted between the equilibration
time for specimen with and without UVGI exposure.
Fig 2 Drying suction-water content SWCC

4.3 Effect of UVGI on Membrane Pore Size
Figure 3a and 3b present typical AFM images of
MWCO 3500 membrane with and without UVGI
exposure at the end of the osmotic tests. The
membrane is composed of series of well-organized
regenerated cellulose acetate strands to form a meshlike formation. AFM images revealed that both
membrane pore sizes were larger than that specified
by the manufacturer. It is believed that thesedifferences were caused by degradation caused bycellulose degrading microbes present within the soil.
Similarly, degradation also occurred due to UVGI
exposure.

Fig 1 Time equilibration plot in osmotic test at
suction 3.65 MPa.
The specimen that has undergone UVGI was foundto equilibrate much longer (i.e. 8 days) as comparedto 6 days for the specimen without UVGI exposure.
Interestingly, the final water content attained for the
specimen exposed under UVGI was slightly greatercompared to the specimen that has not been exposed
to UVGI. It is expected that a much fasterequilibration was obtained due to alteration of theSome differences were noted between the poresize of membrane after the test and the pore size ofmembrane exposed to UVGI. Differences in the
structural arrangements of the strands were also
observed. Alteration of the membrane pore sizes bycellulose degrading microbes in this study (i.e.
218

GEOMATE- Bangkok, Nov. 14-16, 2016

without UVGI) was found to be greater than thosethat have undergone UVGI exposure. Surprisingly,
although alteration of the membrane pore sizes
occurred after UVGI exposure, the final watercontents determined in the osmotic tests were foundto be in better agreement with those of the osmotic
test at lower suctions and the vapour equilibrium
water contents.

Slight differences were noted on 3369 cm-1 peaks
corresponding to O-H bond. It is noted that a slightly
higher transmittance were obtained for PEG solution
after osmotic tests without UVGI exposure. Apartfrom that, the primary molecular structure of PEG
remained somewhat unaffected by the exposure of
UVGI, indicating that no degradation occurred.

5.0 CONCLUSION
Results of the investigation after conducting
osmotic test under UVGI exposure have beenpresented in this paper. Based on the test results, thefollowing conclusions were drawn:i.
All soil microbes in this study werecompletely eliminated by 10 minutes of
UVGI exposure, and thus degradation of
membrane by microbes is prevented.
ii.
Water contents obtained after UVGIcomplemented vapour equilibrium andosmotic tests results at lower suctions, andmore precise SWCC were established.
iii.
the FTiR Spectrum showed that UVGI had
some effects on the O-H bond and pore sizeof the cellulose acetate semipermeable
membrane. These effects were found to be
insignificant in the determination of final
water contents of the bentonite.
ACKNOWLEDGEMENTS
The authors would like to express their gratitude
to Universiti Malaysia Pahang for providing
financial support under RDU 130384. The authors
would also like to thank the Malaysian Nuclear
Agency for providing AFM service for this study.

Fig. 3 AFM images of MWCO 3500 semipermeable membrane after osmotic tests (a) without
UVGI and (b) with UVGI exposure
4.4 Effect of UVGI on FTiR Spectrum

REFERENCES

Comparison of FTiR spectrums for PEG 6000
solutions after completion of osmotic tests with andwithout UVGI exposure is shown in Fig. 3.

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220Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STABILITY ANALYSIS OF SLOPES ON CLAY SOIL FOUNDATIONS BY
LIMIT EQUILIBRIUM AND FINITE ELEMENT ANALYSIS METHODS
Abdul Karim M. Zein1 and Waleed Abdul Karim2
Building and Road Research Institute, University of Khartoum, Khartoum, Sudan
ABSTRACTThis paper presents an evaluation and comparative study on the stability analysis of simple slope models foundedon undrained clay soils. The analysis was performed according to the limit equilibrium (LE) and finite element (FE)methods utilizing “Slide 2D” and “Plaxis 2D” computer programs respectively. Forty five slope models withdifferent geometries based on soft, medium stiff and very stiff clay soil foundations were considered for stability
analysis. The comparison made between four LE methods indicated that the Bishop and Spencer methods produced
practically similar FOS results whereas the Fellenius and Janbu methods gave FOS values lower than the Spencer
method by 3.0% and 7.4% respectively. For slopes founded on soft clays, the difference in FOS computed by the LE
and FE methods is negligible but the FOS values computed by the LE methods were 8.5% higher than the FE
method for slopes on medium and very stiff clays. Based on the results of linear regression analysis of all data, theFellenius, Bishop, and Spencer methods gave FOS values higher than the FE method by 1.4, 5.3 and 5.7%
respectively whereas the Janbu method gave FOS lower than the FE method by 4.3%. Evaluation of the effects ofslope geometry and foundation soil properties on slope stability revealed a reliable relationship between FOS and avariable combining four slope and soil parameters.
Keywords: Slope Stability Analysis, Factor of Safety, Undrained Clay Soils, Limit Equilibrium, Finite Element

visualize soil deformations in place. Many design
computer software packages have been developed toenhance the application of the LE and FE methods inslope stability analysis.
This paper presents an evaluation and comparisonof the results of slope stability analysis carried out
according to certain LE and FE methods for a simple
slope with variable geometries founded on undrained
clay soils with different strength characteristics. The
analysis was performed utilizing the “Slide” and“Plaxis” computer software programs for the LE andthe FE approaches respectively. Some important
aspects are given below on the LE and FE slope
stability analysis methods.

INTRODUCTION
Slope stability problems are normally encounteredwhen the balance of natural or engineered soil slopes
is disrupted. The stability of a slope may be definedas the resistance of an inclined surface within the soil
mass to failure by sliding, overturning or collapsing.
Soil stability analysis is performed to assess
equilibrium conditions and achieve safe design ofslopes. The factor of safety (FOS) against slope
instability is defined as the ratio of the resisting
forces (strength) to the driving forces (loading) along
a potential failure surface. The slope is said to be in alimit equilibrium condition when FOS = 1 whereas
higher values correspond to stable slopes.
Slope stability analysis can be evaluated by thelimit equilibrium (LE) and finite element (FE)methods. The LE methods are based on the static
equilibrium of the forces and/or moments whereas
the FE methods utilize the constitutive law or stressstrain relationship concept. The LE approach
involves different methods depending on the type ofproblem to be solved and the required accuracy ofanalysis results. More attention has been directed in
the recent decades to the use of the FE in slope
stability analysis to model cases with complex slope
geometry, soil behavior and loading sequences and to

METHODS OF SLOPE STABILITY ANALYSIS
In the LE slope stability analysis, the soil mass
above the slip surface is divided into slices and theshear and normal inter-slice forces are determined byapplying appropriate forces and/or moment
equilibrium equations for each slice to satisfy thestatic equilibrium conditions. The FOS computation
procedure involves comparing the available soil shear
strength along the sliding surface with the force
required to maintain the slope in equilibrium. The LE
methods are commonly adopted in routine design due

221GEOMATE- Bangkok, Nov. 14-16, 2016

whereas the lower foundation layer is a homogeneous
clay of variable shear strength. The clay soils weremodeled in an undrained condition with zero internal
friction angle ( =0) and cohesion values of 20, 40
and 140kN/m2 representing soft, medium stiff andvery stiff soils. Table 1 lists the input parameters
used for stability analysis which include theproperties of the slope material and foundation soil
types. These properties are typical of those normallyobtained from testing of such soil materials.

to their simplicity, accuracy and the small number ofparameters required for analysis. The first LE
method, known as the Fellenius method or theordinary method of slices, was developed in 1927 by
the Swedish engineers and is applicable to circularslip surfaces [1]. A revised method of circular slip
analysis was developed by Bishop in 1955 [2] for
improving the accuracy of FOS computations. The
latter requires an iterative procedure to calculate theminimum FOS. To undertake analysis of non-circularslips, Janbu’s method [3] is normally used. Morerefined LE methods which account for both force andmoment equilibrium [4] have been developed tofurther improve the FOS calculation accuracy.
Finite Element Stability Analysis Methods
The main feature of the FE is using soil stressstrain behavior for slope stability modeling whichchanges the problem from a static-indeterminate to astatic-determinate one. In FE methods there is no preassumption about the shape and location of the
failure surface. FE modeling starts by dividing theslope into a finite number of zones or elements.
Forces and strains are then calculated for eachelement using the appropriate constitutive laws forthe materials in the slope.
The FE method uses the strength reduction
method (SRM), to calculate/simulate failure limit
state of slope and safety factor. The SRM is based onprogressive reduction of soil strength parameters;
and c, by a “strength reduction factor” (SRF) untilconvergence occurs within a specified number of
iterations and tolerance or failure of slope occurs.
The SRF which corresponds to the point at the lastconvergence state is equivalent to the safety factor.

Fig. 1 Two soil layers model assumed for slope
stability analysis.
Table 1 Properties of the slope material andfoundation soil types used for stability analysis
Soil properties
Slope material
Soft clay
Base
Med.
Soil
stiff clay
Very
stiff clay

MODELLING OF SLOPE STABILITY
PROBLEM

30
0

γb
kN/
m3
18.0
17.3

E
MP
a
15
14

0.17
0.30

40

0

18.9

33

0.30

140

0

20.4

75

0.30

cu
kN/
m2
10
20

Φu
(°)

ν

Simple slope geometries of variable heights (H)and inclination angles (β) were assumed such thatdifferent slope models are covered in the analysis.
Slope heights of 12m, 20m and 28m were chosen to
represent low, medium and high slopes and for eachcase the angle β was varied from 10° to 24°. In total,
forty five study cases were considered for analysis.

A description of the methodology followed formodeling the various aspects of slope stability
problem analyzed is given hereunder. The main
aspects included the slope geometry, characteristics
of slope and foundation soils and the computer
programs used for FOS computations.
Slope Geometry and Soil Properties

Slope Stability Analysis Computer Software Used

The slope geometry assumed in this study iscomprised of a two soil layers model with variable
strength parameters as schematically illustrated inFig. 1. The upper slope layer comprises a sandy soil

Limit Equilibrium “Slide 2D” Software
The LE methods chosen in this study included theFellenius’s (FM), Bishop’s method (BM), Spencer's

222

GEOMATE- Bangkok, Nov. 14-16, 2016

values of FOS less than 1 in all cases indicating thatthe slopes are unstable due to their inadequate shear
strength. The analysis results showed that the slopes
were even more unsafe for the cases with higher H
and β values. The FOS values in Table 2 may be
useful when it is required to improve the resistance ofsuch weak soils against slope failures throughapplication of certain soil stabilization techniques.
The FOS pertaining to slopes on medium stiffclay foundations (Table 3) varied from 0.45 to 1.15
with the lower and upper values pertaining to
inclination angles of 24° and 10° respectively. Such aslope foundation represents a marginal stability
situation for the analyzed slope models. The results
indicated that stability could be achieved (FOS ≥1)
by some methods only in slopes with small heights(H = 12-14m) and mild inclinations (β = 12-14°).
For slopes on very stiff clay foundation soils, theFOS (Table 4) varied from 1.23 to 3.7 indicating that
all analysed models were stable even for the cases ofmaximum H and β values. The highest and lowest
FOS values pertain to slope geometries with thelowest and highest values of H and β respectively.

method (SM) and Janbu's method (JM). The input
and features of models included the slope geometry
and soils parameters, selection of the number ofslices (25 slices) and adoption of the Mohr-Coulomb
soil shear strength model without tension cracks. The
slope models were analyzed using “Slide 2D”
computer Software from Rocscience Inc. [5] forcomputing FOS using vertical slice limit equilibrium
methods of circular or non-circular failure slip
surfaces. The Slide software is simple to use, and yetcomplex models can be created and analyzed quickly
and easily. Searching of the critical slip surface isrealized with the help of a grid or as a slope search inuser-defined area. Individual slip surfaces can beanalyzed, or search can be applied to locate thecritical slip surface for a given slope.
Finite Element “Plaxis 2D” Software
The stability of the slope was analyzed using thefinite element Plaxis 2D software [6] normally usedwith plane strain model for stability and deformation
analysis to determine the minimum FOS for assumed
slope models. The models were drawn with 15 nodes
and the standard fixities were used to define theboundary conditions. Plaxis 2D uses a convenientgraphical user interface that enables users to quickly
generate a geometry model and finite element meshbased on a representative vertical cross section of thesituation in question. Standard boundary conditions
are automatically generated by the program. All soil
types are modeled through Mohr-Columb yield
criterion in Plaxis. Once the geometry model andboundary conditions are defined, automatic meshgeneration is applied with the bandwidth optimizer
for the finite element discretization refinement.

Comparisons of FOS Computed by LE and FE
Methods
GeneralFor all slopes founded on soft clays there is a very
good agreement between the FOS values computed
according to the LE methods except for the JM whichgave consistently low FOS with a discrepancy
reaching 10% for the 12m high slope. Generally, theFE method indicated FOS values that are similar to orslightly lower than those of the LE methods for
slopes founded on these soils. Therefore, both
approaches seem to produce similar FOS values.
For slopes founded on medium stiff clays, there isa perfect agreement between the FOS computed by
the BM and SM methods; both methods gave thehighest FOS followed by the FM whereas the JM
gave the lowest values. The FE method showed areasonable comparison with the LE methods for
slopes of low to moderate heights but gavesignificantly lower FOS for relatively high slopes.
As noted for the slopes founded on soft andmedium stiff clays, the BM and SM gave the highest
FOS values for very stiff clays, followed by the FM
whereas among the LE methods the JM indicated thelowest values. There is a very good agreement
between the FOS values deduced from the FE

RESULTS AND DISCUSSION
Stability Analysis Results
The results of FOS computed according to the LE
and FE stability analysis methods are presented inTables 2, 3 and 4 for the slopes founded on soft,
medium stiff and very stiff clays respectively. The
discussion presented herein is focused oncomparisons of FOS computed using LE and FE
methods and evaluation of the effects of slope
geometry and soil properties on slope stability.
For slopes founded on soft clays it may be notedfrom Table 2, that the LE and FE methods yielded223

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 Computed FOS values for slope models founded on soft clay soil foundations
Slope
Geometry

H

12m

20m

28m

β

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

BM

0.60

0.56

0.54

0.54

0.53

0.46

0.39

0.36

0.35

0.34

045

0.33

0.29

0.28

0.26

SM

0.60

0.56

0.54

0.54

0.53

0.46

0.39

0.36

0.35

0.34

0.45

0.33

0.29

0.28

0.27

FM

0.59

0.56

0.54

0.54

0.53

0.47

0.39

0.36

0.36

0.34

0.46

0.36

0.31

0.30

0.28

JM

0.55

0.51

0.50

0.50

0.50

0.43

0.36

0.33

0.32

0.31

0.42

0.31

0.27

0.26

0.25

FEM Plaxis 2D

0.61

0.57

0.54

0.53

0.52

0.45

0.38

0.36

0.35

0.33

0.39

0.36

0.30

0.28

0.26

Limit
Equilibrium
Methods
Slide 2D

Table 3 Computed FOS values for slope models founded on medium stiff clay soil foundations
Slope
Geometry
Limit
Equilibrium
Methods
Slide 2D

H

12m

20m

28m

β

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

BM

1.16

1.09

1.07

1.06

1.05

0.84

0.84

0.69

0.68

0.66

0.73

0.77

0.55

0.53

0.50

SM

1.16

1.09

1.06

1.06

1.04

0.84

0.73

0.68

0.67

0.66

0.77

0.61

0.55

0.53

0.50

FM

1.12

1.06

1.03

1.03

1.01

0.81

0.70

0.66

0.65

0.63

0.73

0.59

0.52

0.51

0.49

JM

1.05

0.99

0.97

0.96

0.95

0.77

0.67

0.62

0.61

0.60

0.72

0.56

0.50

0.48

0.46

FEM Plaxis 2D

1.09

1.01

1.04

1.03

1.01

0.75

0.66

0.62

0.61

0.59

0.60

0.51

0.47

0.45

0.43

Table 4 Computed FOS values for slope models founded on very stiff clay soil foundations
Slope
Geometry

H

12m

20m

28m

β

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

10°

14°

18°

20°

24°

BM

3.85

3.24

2.60

2.37

2.02

2.61

2.34

2.24

2.13

1.80

2.16

1.83

1.70

1.66

1.61

SM

3.84

3.23

2.60

3.37

2.02

2.60

2.33

2.22

2.12

1.80

2.14

1.81

1.68

1.64

1.60

FM

3.76

3.12

2.49

2.27

1.94

2.51

2.23

2.12

2.05

1.73

2.04

1.71

1.58

1.54

1.48

JM

3.38

3.10

2.46

2.24

1.92

2.32

2.04

1.94

1.91

1.72

1.93

1.60

1.47

1.43

1.38

FEM Plaxis 2D

3.82

3.31

2.60

1.76

1.61

2.62

2.34

2.19

1.52

1.33

2.14

1.83

1.68

1.39

1.23

Limit
Equilibrium
Methods
Slide 2D

method on one hand and the BM and SM methods on
the other for slopes of low to moderate inclinations (β=12 to 18°). For slopes with β =20°and steeper, theFE method gave FOS significantly lower than the LE
methods. The differences in FOS computed by the
LE and FE methods become more pronounced forslopes with relatively steep inclinations (β =24°) andgreater heights (H >20m).

The SM was selected as a basis for comparison
because it satisfies both the force and moment
requirements for static equilibrium. Moreover, theSM can be applied for circular as well as non-circularslip surfaces [4]. The comparison criterion was thediscrepancies between FOS values deduced from the
SM and other methods. The LE-LE comparison
results show that the FOS values computed by the
BM and the SM are practically the same for all slopemodels and foundation conditions. The FOS values
computed using the FM were on average 3% lower
than SM values for slopes on clays with different
strength characteristics. The JM gave FOS values
lower than the SM by 4.5 to 9.1% with an average of
7.4%. For the medium and very stiff clay

LE – LE Comparisons
In the LE-LE comparison, the FOS values computedfrom the Spencer method (SM) were compared tothose obtained from by the other three LE methods.

224

GEOMATE- Bangkok, Nov. 14-16, 2016
International Conference on Geotechnique, C nstruction Materials
nvironment, Bangkok, Thailand
ISBN: 8foundations, the discrepancy between the JM and SM
methods were higher than the previously reported
maximum difference of 6% [7] between various LE
methods.

Prediction of FOS for Simple Slopes Founded on
Undrained Clay Soils
The effects of slope geometry and foundation soil
characteristics on the computed FOS were evaluatedusing the data pertaining to the LE and FE slope
stability methods. The parameters considered forevaluation included the slope height (H) andinclination angle (β) and the undrained cohesion (c)and bulk unit weight (γ)γ) of foundation soils. For agiven slope the stability increases with soil strengthand unit weight (c and γ) and deceases with slope
height and inclination angle (H and β). Hence, the
FOS may be expressed as a function of slope and soil
parameters by the following equation:

LE-FE Comparisons
The average FOS values obtained for the LE
methods for each slope model were compared to thevalues deduced from the FE method. The results
indicated that the FE method gives FOS values which
in all cases are smaller than the average LE values.
The difference in FOS computed by the LE and FE
methods is negligible (0.9%) for slope models
founded on soft clays. However, the discrepancy
between the two approaches becomes more
pronounced (8.5%) for slopes founded on medium
and very stiff clays.
To examine whether the FOS computed from the
LE methods can be related to the FOS according to
the FE method the two data sets were analyzed by the
regression method. Four linear equations were
established between the FOS data pertaining to each
LE method and the LE method. Equations of the best
fit lines indicated that the Fellenius, Spencer andBishop methods gave FOS higher than the FE
method by 1.4, 5.3 and 5.7% respectively whereas
the Janbu method gave FOS lower than the FE
method by 4.3%. The coefficient of determination R2
varied from 0.958 to 0.967 indicating a perfect linear
relationship between the LE and FE methods .
It has been reported in a previous study [8] thatthe differences in FOS computed by the FE and LE
are negligible for simple slopes founded on undrained
clay soils. The results of this study reveal that such afinding is only applicable for the slopes founded on
soft clays. For slopes founded on medium and very
stiff clays the FEM tends to give FOS values lower
than the LE methods. The difference may be
attributed to the degree of accuracy of the inter-slice
forces calculation [9]. The inter-slice forces are moreaccurately calculated by the FE software packages
which take into account the local stress distribution in
the soil mass. On the other hand, the LE methods
have limitations with regards to the inter-slice shear
forces computation. It has been indicated that theforces computed in the FE method are higher than in
the LE methods; thus lower FOS are produced by theformer.

∗

(1)

∗

FOS

To define the type of relationship in above
equation, the FOS values computed by each method
were plotted against the results of the product of(γ*c/H*tan β) as shown in Fig. 2 with H, c and γ
expressed in m, kN/m2 and kN/m3units respectively
4.5
4.0
3.5
3.0
2.5
2.0
1.5
1.0
0.5
0.0
0

200

400

FM

BM

JM

SM

600

800

1000

FEM

1200

1400

(γ*c/H*tanβ)β)Fig. 2 Relationship between FOS and (γ*c/H*tan(
The trends depicted in Fig. 2 indicate that a soundlinear relationship exists between the FOS and(γ*c/H*tan β) variables for each analysis method.
Linear regression analysis was carried out using data
pertaining to different slope and foundation
conditions to quantify such relationships. The
following general equation was developed for
estimating the FOS from slope and soil parameters:
∗
∗

(2)

For each method of analysis, the constants A and
B are dependent on the slope geometry and soil

225

GEOMATE- Bangkok, Nov. 14-16, 2016
The effects of soil geometry and foundation soil
properties on the FOS computed according the twostability analysis approaches were evaluated. A
simple and reliable mathematical linear relationshipwas developed from regression analysis of data
between FOS and a variable combining fourparameters of slope geometry and soil properties asgiven in Eq. 2. The constants A and B in thisequation are dependent on slope geometry, soil
properties and applied method of analysis.
Finally, the study results indicate that the LE
slope stability analysis methods which are simple andrelatively fast can produce accurate and reliable
results. Such methods may be applied withconfidence in routine design for slopes with simple
geometry founded on undrained clays. The FE
method is an advanced and reliable analysis tool ofgreat value in modeling design cases with complex
slope geometry, heterogeneous soil behavior anddifferent loading patterns.

properties and they respectively represent the slope
and y-intercept of the best fit lines for the data setsplotted in Fig. 2. Their values were determined foreach method as given in Table 5. The data in Table 5
confirm the previously stated conclusion that the FE
method tends to give FOS that are lower than all LE
methods (particularly the BM and SM).
Equation 2 may be used in conjunction with thevalues of A and B in Table 5 to predict the FOS using
the LE and FE methods for slopes of simple
geometry founded on homogeneous clay soils withsimilar characteristics.

Table 5 Values of A, B and R2
BR2

Analysis MethodALE
methods
based onSlide
softwareBM

0.003 0.361

0.902

SM

0.003 0.361

0.904

FM

0.002 0.353

0.913

JM

0.002 0.336

0.903

FE Plaxis 2D method

0.002 0.317

0.921

REFERENCES[1] Fellenius, W. “Erdstatische Berechnungen mit
Reibung and Kohaesion”, Ernst, Berlin, 1927.
[2] Bishop, A.W. “The Use of the Slip Circle in the
Stability Analysis of Slopes”. Géotechnique, ,Vol. 5, No. 1, 1955, pp. 7-17
[3] Janbu, N. “Slope stability computations”, Soil

CONCLUSIONS

Mech. Found. Engng Report, Trondheim:
Technical University of Norway, 1968.
[4] Spencer, E. “A Method of Analysis of the

The stability of simple slope models founded onclay soils with variable undrained strengthcharacteristics was analyzed using the LE and the FE
methods utilizing the “Slide” and “Plaxis” software
programs respectively. The main conclusions drawn
from analysis and discussion of study results are
given below.
The results of comparison between the FOS
computed using four LE analysis methods revealed
that the Bishop and Spencer LE methods produced
similar FOS values; whereas the Fellenius and Janbu
methods gave FOS lower than the Spencer’s methodby 3.0% and 7.4% respectively.
The FOS computed using LE and FE methods
compared very well for the slope models founded on
soft clays. However for slopes on medium and very
stiff clay foundations, the LE methods tend to give
FOS values higher than the FE method.
The results of linear regression analysis of data
deduced from the LE and FE methods showed that
the FOS obtained from Bishop, Spencer andFellenius LE methods are higher by 1.4 to 5.7%
compared to those from the FE method.

Stability of Embankments Assuming Parallel
Interslice Forces” Géotechnique, Vol. 17, No. 1,
1967, pp. 11-26[5] Rocscience Inc., 2008, SLIDE version 5.0 User
Manual[6] PLAXIS, “Finite Element Code for Soil and RockAnalysis,” PLAXIS-2D Version 8, 2004, Manual,
DUT, Netherlands,[7] Duncan, J., M. State of the art: Limit equilibrium
and finite element analysis of slopes. J.of
Geotech. Engng., Vol. 122 (7), 1996, 577-596.
[8] Khabbaz, B., Fatahi, B. and Nucifora C “Finite
Element Methods against Limit Equilibrium
Approaches for Slope Stability Analysis” Proc.
ANZ Conf. , 2012, pp. 1293-1298.
[9] Aryal, K.P., Differences between LE and FE
Methods used in Slope Stability Evaluations, The
12th Int. Conf. of Int. Association for Computer
Methods and Advances in Geomechanics , 2008.

226

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EFFECT OF PERIODICAL RAINFALL ON SHALLOW SLOPEFAILURES: FINITE ELEMENT ANALYSIS
Avirut Chinkulkijniwat1, Somjai Yubonchit2, Rajeshwar Goodary 3 and Duc Bui Van4
School of Civil Engineering, Suranaree University of Technology, Nakhon Ratchasima, Thailand.
3
Faculty of Sustainable Development and Engineering, Université des Mascareignes, Mauritius.

1,2,4

ABSTRACT
Assessment of rainfall-induced shallow slope failures is vital to mitigate damages of infrastructures, and lossof people life in the hazardous areas. The usual tool for prediction of shallow slope failures is critical rainfall
concept due to its simplicity and faster assessment. However, this concept has been developed based on statistical
approach from historical slope failure data, in which many factors those might trigger shallow slope failures are
disregarded. Assessment of the probable factors triggering shallow slope failures is presented in this paper. A setof parametric study was numerically conducted under finite element environment. The rainfall periods to slope
failures (𝐷𝑓 ) under various rainfall intensity (𝐼𝑓 ) were recorded. The relationships between 𝐼𝑓 and 𝐷𝑓 for initiationof shallow slope failures (ID thresholds) were presented. The results indicate that periodical rainfall plays role on
failure state of shallow slope. The 𝐷𝑓 under assigned 𝐼𝑓 decreases with increasing dry period.
Keywords: Periodical rainfall, Shallow slope, ID thresholds

they conducted stability analysis only under a specificperiod of rainfall. None of them performed theanalysis with continuous rainfall until the slope
failure takes place. Set of parametric study were
performed via finite element based software toevaluate the influential factors on the shallow slope
failure. The factor focused in this study was
periodical condition of rainfall. The slope failure
were triggered by assigned continuous rainfall withvarious rainfall intensities. Times to failure according
to the assigned rainfalls were recorded and the ID
thresholds were plotted accordingly.

INTRODUCTION
A critical rainfall concept is a common tool for
mitigating the slope failure worldwide. This concept
has been developed based on a statistical approach
related to historical slope failure data, and shown inform of rainfall thresholds that trigger the failures (IDthresholds). It is presented by a relationship between
rainfall intensity and duration at the state of slope
failure [1]-[8]. The advantages of critical rainfall
concept are their simplicity and fast assessment. It istherefore implemented as a part of early warning
system in various countries, for example in Hong
Kong [9], USA [10], [11] and Italy [12]-[14].
Although the critical rainfall concept has beenused worldwide, possible factors triggering the slope
failures have been disregarded [15]. Rahardjo et al.
[16] performed a set of parametric study using
numerical model to investigate influential factors on
instability of homogenous soil slopes. The results
indicated that the stability of homogenous soil slope
or safety factor (FS) is affected by many factors
including saturated permeability of soil, rainfall
intensity, slope angle, and also initial water tablelocation. Rahimi et al. [17] studied effect of hydraulicproperties on instability of slope subjected to rainfall
events. They stated that the soil-water characteristic(SWC) play an important role on FS especially inpoor drainage soil. While the saturated coefficient of
permeability play a vital role on the stability ofhomogenous slope both for good and poor drainage
soil slopes. In addition, the effect of antecedent
rainfall on instability of soil slopes has been
examined [18], [19].
Previous researches have recognized that manyfactors could trigger instability of slopes. However,

THEORETICAL BACKGROUND
Finite element PLAXIS code with a fully coupled
flow-deformation analysis was used in this study. The
code to rainfall infiltration problems was verified by
Hamdhan and Schweiger [20]. Mohr-Coulomp
failure criterion with Bishop’s effective stress [21] forunsaturated soil is used:𝜏 = 𝑐 ′ + (𝜎𝑛 − 𝑢𝑤 )𝑡𝑎𝑛𝜑′ + 𝜒(𝑢𝑎 − 𝑢𝑤 )𝑡𝑎𝑛𝜑′ (1)where 𝜏 is shear strength of unsaturated soil, 𝜎𝑛 istotal normal stress, 𝑢𝑎 is pore air pressure, 𝑢𝑤 ispore-water pressure, (𝜎𝑛 − 𝑢𝑎 ) is net normal stress,(𝑢𝑎 − 𝑢𝑤 ) is metric suction, 𝑐′ is effective cohesion,
𝜑 ′ is internal soil friction angle and 𝜒 is scalarmultiplier which was assumed equal to the effective
degree of saturation(𝜃𝑒 ).
Richard’s equilibrium was used to simulatetransient flow through unsaturated soil. The
equilibrium states of flow in 𝑥, 𝑦 and 𝑧 directions is:227

GEOMATE- Bangkok, Nov. 14-16, 2016

𝜕ℎ
𝜕ℎ
𝜕
𝜕
(𝑘 (ℎ) ) +(𝑘 (ℎ) )
𝜕𝑥
𝜕𝑦
𝜕𝑥 𝑥
𝜕𝑦 𝑦
𝜕ℎ
𝜕
+ (𝑘𝑧 (ℎ)
+ 1)
𝜕𝑧
𝜕𝑧
𝜕ℎ
= {𝐶(ℎ) + 𝑆. 𝑆𝑠 }𝜕𝑡(2)where 𝑘𝑥 , 𝑘𝑦 , 𝑘𝑧 are coefficients of permeability indirections of 𝑥, 𝑦 and 𝑧 respectively, 𝐶(ℎ) =(𝜕𝜃 ⁄𝜕ℎ) is differential between volumetric moisture
content(𝜃) to pressure head(ℎ) which represent the
specific moisture capacity, and 𝑆𝑠 is the specificstorage of a material.
Permeability of unsaturated soil depends mainly
on soil-water characteristics (SWC). The SWC is therelationship between moisture content and pressure
head, which can be explained by van genuchten [22]and the permeability function is explained by Mualem[23] as respectively shown as Eq. 3 and Eq. 4.
𝜃𝑒 =𝜃𝑊 −𝜃𝑟𝑒𝑠𝜃𝑠𝑎𝑡 −𝜃𝑟𝑒𝑠1= [1+(𝛼ℎ)𝑛]𝑚
1

(3)
1

𝑚 2

𝑘(ℎ) = 𝑘𝑠𝑎𝑡 {𝜃𝑒 ⁄2 [1 − (1 − 𝜃𝑒 ⁄𝑚 ) ] }(4)Previous literatures [15], [16] reveal that thesaturated permeability plays a major role on thestability of slope. Hence, it was mainly focused in thispaper. The remained parameters were kept constant at
𝑐 ′ = 6.74 kN/m2, 𝜑 ′ =33.6˚, 𝛼 = 0.162 kPa-1 and n =1.564. These parameter values were deducted fromthe average values reported in Table 1. In this study,
variation of the saturated permeability is represented
by type of soil, i.e. the soils A, B and C stand for low(𝐾𝑠𝑎𝑡 =1x10-6 m/s), medium (𝐾𝑠𝑎𝑡 =1x10-5 m/s) andhigh (𝐾𝑠𝑎𝑡 =1x10-4 m/s) drain ability, respectively.
Eighteen cases of simulation run were conducted
to evaluate the effect of periodical rainfall on time to
failures of shallow slope. Rainfalls assigned in thispaper were composed of two rainfall events having
the equal rainfall intensity. The first rainfalls wereassigned for 24 hr followed by dry period and thesecond rainfall which were assigned until the slope
failure. The durations of dry period taken place
between rainfall events (𝑡𝑏 ) were 2 and 7 days. In thisstudy, the effect of time between rainfall events isevaluated under three rainfall intensities and two
types of soil, while slope angle was also kept constantas summarized in table 2.
SET UP OF EXPERIMENTS

where 𝜃𝑤 is volumetric moisture content, 𝜃𝑟𝑒𝑠 isresidual volumetric moisture content, 𝜃𝑠𝑎𝑡 issaturated volumetric moisture content, 𝑘𝑠𝑎𝑡 issaturated permeability of soil and 𝛼 are related to airentry value of soil, which is the matric suction valuethat must be exceeded before air recedes into the soil
pores, 𝑛 is the rate of water extraction from the soil
once the air entry has been exceed, respectively, , 𝑚
is parameter related to value of 𝑛(𝑚 = 𝑛 − 1⁄𝑛).
These two group of material parameters including
shear strength parameters ( 𝑖. 𝑒. 𝑐 ′ , 𝜑′ ) and three
hydraulic related parameters (𝑖. 𝑒. 𝛼, 𝑛, 𝑘𝑠𝑎𝑡 ) are therequired parameters to perform an analysis of rainfallinduced slope failures in PLAXIS. In this study, theseparameters were obtained by pervious research works
and discussed in the following section.

Slope geometry, boundary conditions and fixity usedin this paper are shown in Figure 1. The modelcomposed of two layers: natural soil and bedrock.
Bedrock layer is overlaid by soil layer of 3 meters
thickness, which gives a soil depth over slope lengthof 3/100. Standard fixities were prescribed to allow
only vertical movement along the boundary sides,
while lateral and vertical movements were fixed atbottom boundary. Fifteen-node triangles finite
element mesh was assigned. The finer elements was
generated at the soil layer, and the finest mesh was
generated at interested zone.
A prescribed flux, which is used to represent thedesired intensity of rainfall, was assigned along theslope surface BC. Based on the rainfall intensity
specified in Table 2, a range of pore pressure was
between -0.05 m and 0.05 m. By this maximum porewater pressure of 0.05 m, the pounding water due tothe excess of rainfall intensity over the drainage
capacity at soil saturation state could be developed up
to 5 cm. over the slope surface. While the minimum
pore water pressure of -0.05 m was used to represent
a depth where the negative flux due to evaporation is
no longer occur. The boundaries AB and CD wereassigned as no flux boundary, while the boundaries
AHG, DEF and GF were prescribed as impervious
boundary. The initial conditions of the model isaccording to no groundwater table exists in the soil
layer. Table 3 summarizes parameters required for
calculation.

MATERIALS AND METHODS
Shallow slope failure is common found in many partsof the world. The geological conditions in eachhazardous area are different depending on climate
conditions, soil type, slope geometry, etc. This paper
gathered the soil properties reported from the relevant
literatures [24]-[30] and they are summarized inTable 1. Typical range of the parameters are as
following: the parameter 𝛼 is between 0.016 and
0.360 kPa-1, the desaturation parameter 𝑛 is between
1.290 and 2.780, 𝜃𝑠𝑎𝑡 is from 0.286 to 0.480, 𝜃𝑟𝑒𝑠 isfrom 0 to 0.250, and the saturated permeability of soil
is between 1.0x10-6 and 2.1x10-4 m/s.

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GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Summary of soil parameters from previous study
Hydrualic property𝑘𝛼Data No.
Author/Year
𝑠𝑎𝑡
(m/s)
(𝑘𝑃𝑎)−1
1
2
3
4
5
6
7
8
9

Jotisankasa et al.(2008) &Jotisankasa A. andMairaing W.(2010)
Bordoni et al.(2015)
Dahal et al.(2009)
Oh H. and Lu N.(2015)
Vieira et al.(2010)
Godt JW. and MaKenna JP(2008)

MEAN

-

B
1
10

C
1
10

Strength property𝑐′
𝜑′(kN/m2)(°)

2.1x10-4
1.0x10-5
2.0x10-5
4.9x10-5
5.6x10-6
1.0x10-6
5.0x10-5

0.360
0.290
0.265
0.066
0.016
0.044
0.096

1.290
1.316
1.596
1.298
1.300
1.370
2.780

6.5
12.8
8.7
17.6
0
4.9
0
6
4.2

37.0
33.1
38.6
28.7
32.0
31.5
34.1
34
33.6

4.29x10-5

0.162

1.564

6.74

33.6

sharply to critical value after 48 hours. In case of
rainfall intensity equals to saturated permeability(saturated permeability of soil type B is 1x10 -5m/s =10-5 x 103 mm/m x 3600 s/hr =36 mm/hr), the FS
decreases sharply since the beginning of rainfall
period.
Figure 3 shows pore water pressure profile(section a-a in Figure 3) related to rainfall intensity of
10 and 36 mm/hr. For the rainfall intensity of 10
mm/hr, the development of pore water pressure isdivided into two stages. The first stage is according to
the rainfall infiltration process. The pore waterpressure increases from its initial value to about -3
kPa. The stability of slope shown in Figure 2 is still
stable during this process because of the remained
matric suction in the soil slope.
After the first stage, rising of water table and thepositive pore water pressure take place at the second
stage which is taking place when the rain water
infiltrates to soil-bedrock interface. The positive porewater pressure results in the reduction of the soil shear
strength especially at the soil-bedrock interface.

Table 2 Summary of numerical experiments
Rainfall intensity relatedTime between
to soil type 𝑖, (mm/hr)
rainfall events 𝑡𝑏 ,
(day)
A
0.36
3.6

𝑛

2,7

Fig. 1 Slope geometry and boundary conditions
NUMERICAL RESULTS AND DISCUSSIONS
The results from numerical experiments are
represented in this section, including 1) the possible
failure mechanism related to pore-water pressure
responses, and 2) rainfall threshold for the initiationof shallow slope failure.
General Mechanism of Rainfall-induced Shallow
Slope Failures
Two cases of simulation results were used to explain
the shallow slope failure mechanism. Rainfall
intensity of 10 and 36 mm/hr were assigned to soil
type B ( 𝑘𝑠𝑎𝑡 =1x10-5 m/s). Figure 2 shows thevariation of FS with period of the second rainfall. For
the rainfall intensity of 10 mm/hr, a small rate of FS
reduction is found at the beginning. The FS decreases

Fig. 2 Relationship between safety factor and time
under rainfall intensities

229

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 3 Material parameters required in PLAXISParameter
SymbolGeneral parameters
Material modelType of material behavior
Dry unit weight
Total unit weight
Strength parameters
Cohesion
Friction angle
Flow parameters
Unsaturated flow model
Soil type
Saturated permeability of soil𝑛𝛼Soil layer

Bedrock layer

Unit

Model
Type
𝛾𝑢𝑛𝑠𝑎𝑡𝛾𝑠𝑎𝑡Morh Coulomb
Undrained A
17.36
17.36

Morh Coulomb
Non-porous
23a*
23a*

kN/m3
kN/m3

𝑐′
𝜑′

6.74
33.62

25a*
50a*

kPa
°

-

m/s
kPa-1

𝑘𝑠𝑎𝑡,𝑥 = 𝑘𝑠𝑎𝑡,𝑦𝑛𝛼Van Genuchten
AB
C
1x10-6 1x10-5 1x10-4
1.564
0.162

ID threshold, respectively. Figure 4 presents the ID
thresholds plotted from all simulation results. A
Single linear relationship between log( 𝐼𝑓 ) and log( 𝐷𝑓 ) could be drawn regardless the magnitude of
permeability of the soil. The effect of periodical
rainfall on ID thresholds is also presented in Figure 4.
Times between the rainfall events were 2 and 7 days.
The smaller 𝐷𝑓 is encountered for the shorter the dry
period. It is found that the effect of time between
rainfall events results in the vertically shift of the ID
thresholds plot.
Fig. 3 Pore water pressure profile under different
rainfall intensity of soil type B
For the rainfall intensity of equal to saturated
permeability, the development of pore water pressure
increases from its initial value to reach null pore waterpressure immediately since rainfall started. In theother word, the matric suction was dissipated during
rainfall infiltration process. It is therefore resulting indrop of FS and then slope failures as shown in Figure2.
Rainfall Thresholds for Initiation of Shallow Slope
Failure (ID thresholds)

Fig. 4 ID thresholds with varying in time between
rainfall events

Generally, the ID thresholds is established by plotting
the relationship between rainfall intensity 𝐼𝑓 versus
duration 𝐷𝑓 at failure states of slope on double
logarithm relationship, which can be explained as:
𝐼𝑓 = 𝑎 + 𝑐𝐷𝑓𝑚

CONCLUSIONS
1) Under a certain slope geometry, shallow slope
failure can be triggered under either the rainfall
infiltration or the rising of water table modes
depending on the soil properties and rainfall
intensity. The soil saturated permeability is one ofthe primary factored control the range of rainfall
intensity. The lower rainfall intensity comparing(5)

where a, c, and m are model parameters, which
represent the curvature, interception and gradient of230

GEOMATE- Bangkok, Nov. 14-16, 2016

with the saturated permeability, the higher
stability of the slope during infiltration stage. For
rainfall intensity equals to or greater thansaturated permeability, the matric suction willcompletely disappear during infiltration stage,and hence the slope failure is possibly foundduring the infiltration state.
2) The periodical rainfall plays an important role on
failure state of shallow slope. The time to slope
failure (𝐷𝑓 ) subjected to assigned rainfall intensity(𝐼𝑓 ) decreases with a short period between rainfall
events. Furthermore, it also results in the fitting
parameter representing interception of rainfall
intensity-duration for initiation of shallow slope
failures (ID thresholds).

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Geomechanics 13(5):653-658.
[21] Bishop AW, and Blight GE. 1963. Some aspectsof effective stress in saturated and partly
saturated soils. Géotechnique 13(2):177-197.
[22] van Genuchten MT. 1980. A closed-form
equation for predicting the hydraulicconductivity of unsaturated soil. Soil Science
Society of American Journal 44:615–628.
[23] Mualem Y. 1976. A new model predicting thehydraulic conductivity of unsaturated porous
media. Water Resources Research 12:513-522.
[24] Jotisankasa A, and Mairaing W. 2010. SuctionMonitored Direct Shear Testing of Residual
Soils from Landslide-Prone Areas. Journal of
GeotechnicalandGeoenvironmental
Engineering 136(3):533-537.
[25] Jotisankasa A, and Vathananukij H. 2008.
Investigation of soil moisture characteristics oflandslide-prone slopes in Thailand. Proceedingof International Conference on Management ofLandslide Hazard in the Asia-Pacific Region,
Sendai Japan:1-12.

[26] Bordoni M, Meisina C, Valentino R, Bittelli M,
and Chersich S. 2015. Site-specific to localscale shallow landslides triggering zones
assessment using TRIGRS. Natural Hazards andEarth System Science 15(5):1025-1050.
[27] Dahal RK, Hasegawa S, Nonomura A,
Yamanaka M, Masuda T, and Nishino K. 2008.
Failure characteristics of rainfall-induced
shallow landslides in granitic terrains of
Shikoku Island of Japan. Environmental
Geology 56(7):1295-1310.
[28] Oh S, and Lu N. 2015. Slope stability analysis
under unsaturated conditions: Case studies ofrainfall-induced failure of cut slopes.
Engineering Geology 184:96-103.
[29] Vieira BC, Fernandes NF, and Filho OA. 2010.
Shallow landslide prediction in the Serra do Mar,São Paulo, Brazil. Natural Hazards and Earth
System Science 10(9):1829-1837.
[30] Godt JW, and McKenna JP. 2008. Numericalmodeling of rainfall thresholds for shallow
landsliding in the Seattle, Washington, area.
Review in Engineering Geology 20:121-136.

232

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EXPRESSION OF AIR PRESSURE DISTRIBUTION USING
SOIL/WATER/AIR COUPLED ANALYSIS

1

Katsuyuki KAWAI1, Takuya KAWAKATSU2 and Atsushi IIZUKA3
Dept. Of Civil and Environmental Engineering, Kindai University, Japan; 2 Shimizu Corporation, Japan;
3
Research Center for Urban Safety and Security, Kobe University, Japan

ABSTRACT
Recently, the frequency of torrential rainfall has increased due to global climate change, and these events cause
sediment potential failure. It is difficult to predict when and where a slope failure will occur because of theconcentration of heavy rain. Knowing precursory phenomena, however, is effective for disaster reduction.
Nonetheless, some of these phenomena have not been explained in the framework of geotechnical engineering.
Organic smells and strange sounds, known as precursory signs of slope failure, propagate through the atmosphere.
Therefore, it is important to monitor air movement within earth structures. This study focuses on pore air behavior
within the ground due to rainfall infiltration. Here, the infiltration column test combined with monitoring smell,as conducted by Tsuchida et al., was first simulated using the soil/water/air coupled finite element code,
DACSAR-MP. Next, a sloping earth structure exposed to rainfall was simulated. Consequently, it was found that
distribution of pore air pressure was dependent on drainage conditions of air, and that pore air behavior influenced
rainfall infiltration behavior.
Keywords: Unsaturated Soil, Soil/water/air Coupled Analysis, Slope Stability, Rainfall, Air Pressure

INTRODUCTION

Sprinkler
Odor intensity sensor

Recently, torrential rainfall events have tended tooccur locally, and it is difficult to predict damage
from landslides due to rainfall. Therefore, increasing
citizen awareness about disasters is needed to reduce
disaster damage. In this regard, understanding
precursory phenomena for landslides is effective.
There are a lot of successful cases in which
precursory phenomena acted as an alert. Because ofthis, precursory phenomena were introduced onto ahazard map published by Japanese government.
However, some of these phenomena are difficult to
explain theoretically in soil mechanics. In particular,
the generating mechanism of unusual sound and smell
before a landslide has not been clarified. Thesephenomena likely result from rainfall influencing air
within the ground, since both sound and smell aretransmitted by air. In this study, rainfall infiltration
into sloped ground was simulated with soil/water/air
coupled analysis to investigate localization of air
within the ground.

Generation
source of smell

Fig. 1 Rainfall infiltration test by Tsuchida et al.
(%)

10cm deep

100

AIR

Tsuchida et al. conducted soil column testing toinvestigate the relationship between rainfall
infiltration and smell occurrence [1]. The soil column
was installed with a profile moisture meter to measure
the seepage line generated by a sprinkler fitted above
the sandy soil column (Fig. 1). Smell was generated

400
350

20cm deep

80

300

30cm deep

250

60

200
40

150

Odor
intensity

20

100

Odor intensity

Degree of saturation

PRIOR
INVESTIGATIONSOFBEHAVIOR WITHIN THE GROUND

Profile moisture
meter

2cm

50

0

0
0

120

240

360

480

600

720

Rainfall time (min)Fig. 2 Distribution of degree of saturation andodor intensity at the ground surface during
rainfall233

GEOMATE- Bangkok, Nov. 14-16, 2016

Here, nE is a shaping parameter; ε vp is plastic
volumetric strain; M is the stress ratio q p ′ at the′ iscritical state; D is the dilatancy coefficient; psatthe yield stress at saturated state; a and ns are theparameters expressing the yield stress increment due
to desaturation; and λ and κ are compression andexpansion indices, respectively. Darcy’s law isassumed for pore water and air flow as follows.

by a source at the bottom of the soil column, andleakage of smell from the surface of the soil column
was measured with an odor intensity sensor. In thisway, the distribution of the degree of saturation andodor intensity due to rainfall infiltration weremonitored (shown in Fig. 2). First, it was found that
degree of saturation started to increase from thebottom of the column, and the rise of the ground waterlevel was more remarkable than the descent of thewetting front at the given rainfall condition. The odor
intensity at the surface of the column drasticallyincreased after an increase in the degree of saturation
at each depth, a phenomena correlated to the rise of
the groundwater level. Tsuchida et al. concluded thatthe groundwater level, determined by arriving to thebottom of infiltrated rain water, pushed air within the
ground upwards, thus odor generation before alandslide signifies that the groundwater level is rising.
However, their investigation was limited to onedimensional infiltration. In this study, the soil column
test performed by Tsuchida et al. was first simulated,followed by simulation of rainfall infiltration into
virtual sloping ground.

v w =−k w ⋅ gradh(7)v a =−k a ⋅ gradh a , ha =

( )
k =
k k =
(1 − S ) (1 − S ) k
2

m
1
1 

k=
krw k wsat
= Se 2 1 − 1 − Se m  k wsatw
1

The soil/water/air coupled simulation code,
DACSAR-MP [2], was used for rainfall simulation.
The constitutive model for unsaturated soil, proposedby Ohno et al. [3], was adopted and formulated withthe theory of three phase’s mixture material proposed
by Borja [4]. In this way, effective stress is expressed
as follows.
(1)

σ

=
σ − pa 1, ps =
Se s

(2)

S r − S rcs=pa − p w , S e =1 − S rc

(3)

net

a

λ −κ
=
ζ exp (1 − Se ) ns ln a  , MD =
1+ e

1

2

e

e

2m

m

ares

(10)

nSr − S r εv + divv w =0(11)(1 − Sr ) εv + nSr − n (1 − Sr )p a
0 (12)− divv a =pa + p0σ=′ D : ε − C ⋅ Se(13)

Here, D is the elasto-plastic stiffness matrix; ε isthe strain increment tensor; and C is the tensor
expressing change in stiffness due to desaturation.
The soil/water/air coupled problem was formulated
by equations (11), (12) and (13). The soil water
retention characteristic curve (SWRCC), which
demonstrates the relationship between suction andsoil moisture, is dependent on suction history. This

(4)
(5)

0

1
3
1
p′ = σ′ : 1, q = s : s , s =σ′ − p′1 =A : σ′, A =I − 1 ⊗ 1
3
2
3

ares

Here, n is porosity; ε v is volumetric strain; and p0
is atmospheric pressure. The elasto-plastic
constitutive model can be obtained from equation (4)and the force equilibrium equation as follows.

nEp′MD  q +− ε vp 0=ζ p′sat nE  Mp′ 

ra

(9)

Here, krw and krw are the relative permeability ofwater and air, respectively; m is Mualem’scoefficient; and k wsat and k ares are water
permeability at a saturated state and air permeability
at a perfectly dry state, respectively. The continuous
equation of water and air are expressed by theapplication of three phases’ mixture theory.

Here, σ ′ is the effective stress tensor; σ net is the net
stress tensor; 1 is the second-order unit tensor; σ isthe total stress tensor; s is suction; ps is suction
stress; pa is pore air pressure; pw is pore waterpressure; S r is degree of saturation; Se is effective
degree of saturation; and S rc is degree of saturation
at s → ∞ . The yield function is expressed as follows.
′, ζ , ε vp ) MD lnf ( σ=(8)

Here, v w and v a are the velocity of water and air,respectively; k w and k a are permeability of waterand air, respectively; h is total water head; γ w is unit
weight of water; and ha is air pressure head.
Permeability of water and air are expressed by
Mualem’s equation [5] and Van Genuchten’s
equation [6] as follows.

SOIL/WATER/AIR COUPLED ANALYSIS=
σ ′ σ net + ps 1

pa

γw

(6)

234

GEOMATE- Bangkok, Nov. 14-16, 2016

simulation are summarized in Table 1, and the soil
water retention characteristic curves are shown in Fig.
3. Here, a sandy soil commonly used for general earth
structures was assumed. The analytical mesh shownin Fig. 3 was assumed in accordance with Tsuchida etal. The undrained water and undrained air boundaries
were provided for right and left side boundaries. A
flux boundary, corresponding to rainfall intensity,and a drained air boundary ( pa = 0 ) were applied tothe upper boundary. On the bottom of the analyticalarea, two kinds of boundary were investigated. One

hysteresis influences compaction behavior. In thisstudy, the SWRCC model proposed by Kawai et al.
[7] was used.
ONE-DIMENSIONALINFILTRATION SIMULATION

RAINFALL

Tsuchida et al. studied air behavior within groundexposed to relatively small rainfall. In this study,
rainfall intensity large enough to generate a wetting
front was simulated. The soil parameters used for the

Table 1 Soil parameters used for simulation′ (kPa)psat
k wx (m/h)
M

λ

κ

0.087

0.009

1.375

m

a

n

10

1.0

0.8

1500
nE
1.00

0.156
kax (m/h)
15.6

k wy (m/h)

S ri

0.078
kax (m/h)
7.8

0.15
ei
0.7

e
urv
gc
ryi n
ry d
nda
Bou

0.8

w
ry
da
un
Bo

0.6

0.4

rve
cu
i ng
ett

Degree of saturation Sr (%)

1

0.2

0
0.01

0.1

1

10

100

Suction s (kPa)

Fig. 4 Analytical mesh for one dimensional
infiltration simulation

0.6

0.6

0.5

0.5

0.4

0.4

h [m]

h [m]

Fig. 3 Soil water retention characteristic curves

0.3

0.3

0.2

0.2

0.1

0.1

0
0

0.2

0.4

0.6

0.8

0
0

1

0min
30min
60min
90min
120min150min180min240min300min0.0005

Sr

0.001

0.00

pa [kPa]

0.6

0.6

0.5

0.5

0.4

0.4

h [m]

h [m]

(a) Degree of saturation(b) Air pressure
Fig. 5 Rainfall intensity 30mm/h (Case DD)

0.3

0.3

0.2

0.2

0.1

0.1

0
0

0.2

0.4

0.6

0.8

0
0

1

Sr [kPa]

0min
30min
60min
90min
120min150min180min240min300min0.02

0.04

pa [kPa](a) Degree of saturation(b) Air pressure
Fig. 6 Rainfall intensity 60mm/h (Case DD)
235

0.06

0.6

0.6

0.5

0.5

0.4

0.4

h [m]

h [m]

GEOMATE- Bangkok, Nov. 14-16, 2016

0.3

0.3

0.2

0.2

0.1

0.1

0
0

0.2

0.6

0.4

0.8

0min
30min
60min
90min
120min150min180min240min300min

0
0

1

0.01

Sr

0.02

0.03

0.04

pa [kPa]

0.6

0.6

0.5

0.5

0.4

0.4

h [m]

h [m]

(a) Degree of saturation(b) Air pressure
Fig. 7 Rainfall intensity 30mm/h (Case UU)

0.3

0.3

0.2

0.2

0.1

0.1

0

0.2

0.4

0.6

0.8

0min
30min
60min
90min
120min150min180min240min300min

0
0

1

0.05

0.1

0.15

pa [kPa]

Sr(a) Degree of saturation(b) Air pressure
Fig. 8 Rainfall intensity 60mm/h (Case UU)
0.3m

0.6m

Undrained water

0.2m

Drained or undrained air

Undrained air

Drained or undrained water

Undrained water

Undrained air

Drained air

1.2m

Fig. 9 Analytical mesh for rainfall simulation on sloping groundcase, named Case DD, investigated the condition ofdrained air and drained water, using a value of
pw = −5.5(kPa) to indicate an initial distribution ofsuction. The other case, named Case UU, investigatedundrained water and undrained air conditions.
Rainfall intensities of 30 and 60 mm/h were simulated.
Figures 5 and 6 show the distribution of the degreeof saturation and air pressure in Case DD. These
results demonstrate that differences in infiltration
behavior depended on rainfall intensity. Degree ofsaturation of the wetting front, indicating rainfall
infiltration, was 0.8 for a rainfall intensity of 30mm/h,
while it was over 0.9 for a rainfall intensity of
60mm/h. However, the time at which the degree ofsaturation started to increase at each depth was earlier
for the 30mm/h than the 60mm/h rainfall intensity.
This tendency can be explained due to the distributionof air pressure. The descent of the wetting front
compressed air within the ground and increased thedegree of saturation at the surface. Here, entrapmentof air was more remarkable for the 60mm/h rainfall
intensity, and compressed air prevented infiltrated
water from descending. The wetting front for the
30mm/h rainfall intensity case exhibited a 0.8 degreeof saturation, and air was easily exhausted from thesurface. Consequently, air pressure gradually
decreased after peaking at 90 minutes.
Figures 7 and 8 show simulation results obtained
from Case UU. Because the bottom boundary was
undrained water, infiltrated water accumulated here
after the wetting front reached the bottom. Moreover,air pressure monotonically increased. In this case, thehigher air pressure that was generated can predict aburst of compressed air from the ground surface.
RAINFALL INFILTRATION INTO SLOPINGGROUNDOn an actual slope, the direction of groundwater
236

GEOMATE- Bangkok, Nov. 14-16, 2016

0.0

0.5

1.0

0.0

1.0

0.5

0.0

0.5

1.0

0.0

0.5

1.0

(b) 1 hour settling(a) 1.5 hour rainfall
Fig. 10 Degree of saturation (Case DD, 30mm/h)(a) 1.5 hour rainfall(b) 1 hour settlingFig. 11 Degree of saturation (60mm/h, Case DD)(a) 1.5 hour rainfall(b) 1 hour settlingFig. 12 Air pressure (Case DD, 30mm/h)(a) 1.5 hour rainfall(b) 1 hour settlingFig. 13 Air pressure (60mm/h, Case DD)(b) 1 hour settling(a) 1.5 hour rainfall
Fig. 14 Water pressure (Case DD, 30mm/h)(a) 1.5 hour rainfall(b) 1 hour settlingFig. 15 Water pressure (60mm/h, Case DD)

0.0

0.5

1.0

0.0

1.0

0.5

(b) 1 hour settling(a) 1.5 hour rainfall
Fig. 16 Degree of saturation (Case UU, 30mm/h)

0.0

0.5

1.0

0.0

1.0

0.5

(a) 1.5 hour rainfall(b) 1 hour settlingFig. 17 Degree of saturation (60mm/h, Case UU)[kPa](a) 1.5 hour rainfall(b) 1 hour settlingFig. 18 Air pressure (Case UU, 30mm/h)

-7.00

-3.50

1.00

-7.00

-3.50

1.00

(a) 1.5 hour rainfall(b) 1 hour settlingFig. 20 Water pressure (Case UU, 30mm/h)(a) 1.5 hour rainfall(b) 1 hour settlingFig. 19 Air pressure (60mm/h, Case UU)

-7.00

-3.50

1.00

-7.00

-3.50

1.00

(a) 1.5 hour rainfall(b) 1 hour settlingFig. 21 Water pressure (60mm/h, Case UU)
parameters are summarized in Table 1 and Fig. 3.
Rainfall duration was 1.5 hour, and the settling timeafter rainfall is provided for each case.
Figures 10 and 11 show the distribution of thedegree of saturation in the case of a drained bottom
boundary (Case DD) under rainfall intensities of
30mm/h and 60mm/h, respectively. Rainfall of
60mm/h saturated the slope surface after 1.5 hours,
while rainfall of 30mm/h did not saturate the surface

flow depends on slope shape. In this study, rainfall
infiltration into sloping ground was simulated. Figure
9 shows the analytical mesh used in simulations. The
right and left side boundaries were undrained waterand air boundaries. The ground surfaces, including
the crown and the slope, were both flux boundaries
corresponding to rainfall intensity. The drainage
condition of the bottom was changed with onedimensional infiltration simulation.
Input soil
237

GEOMATE- Bangkok, Nov. 14-16, 2016

in the same time period. Infiltrated water
concentrated around the toe of slope during thesettling term. This tendency was more remarkable forthe higher rainfall intensity, shown in Fig. 11. Thiscan be explained from equation (9). Unsaturated
permeability depends on degree of saturation.
Rainfall increases the degree of saturation around theslope surface first, resulting in an increase in thepermeability around the slope surface. Consequently,
flux parallel to slope becomes easier than vertical flux.
Figures 12 and 13 show the distribution of airpressure under rainfall intensities of 30mm/h and
60mm/h, respectively. An high air pressure area first
appeared around the crown under a rainfall intensityof 30mm/h. This was because rainfall caused
downward flux and pushed air toward the bottom asa drained boundary. Figures 14 and 15 show the
distribution of water pressure. Rainfall increased thewater pressure at the surface. However, the highwater pressure area descended during the settlingterm, and negative water pressure appeared at thesurface. This was because air permeability around thesurface decreased due to saturation around the surface,and entrapped air expanded by downward flux. This
led to negative air pressure around the crown. On theother hand, under a rainfall intensity of 60mm/h,
since infiltrated water flowed parallel to the slope, thehighest air pressure appeared around toe of slope.
Next, the same rainfall simulation was conductedunder a condition of an undrained bottom boundary(Case UU). Figures 16 and 17 show the distributionof the degree of saturation under rainfall intensities of
30mm/h and 60mm/h, respectively. Since the bottom
boundary was undrained, water infiltrating from theslope surface flowed towards the toe of slope andaccumulated under the berm, especially under
60mm/h rainfall intensity. During the settling term,
accumulated water spread to the left-hand side.
However, in Case UU, the saturated area from theslope surface due to rainfall was smaller than in Case
DD. This was because air was entrapped more easily
and prevented rainfall from infiltrating. Figures 18
and 19 show the distribution of air pressure. In thesefigures, an area of high air pressure covered the wholeof the sloping ground during rainfall, and it moved
toward the toe of the slope during the settling term.
This tendency was more remarkable and higher airpressure was generated under the higher rainfall
intensity. Figures 20 and 21 show the distribution ofwater pressure. These results show that infiltrated
water reached the bottom and tended to function as
the phreatic surface.

CONCLUSIONSIn this study, rainfall infiltration was simulatedwith a soil/water/air coupled analysis to investigate
rainfall effects on the behavior of air within the
ground. Results show that the geometric condition
and the drainage condition of water air strongly
influenced the distribution of air pressure. Moreover,it was found that high pressure air entrapped by
infiltrated water prevented rainfall from infiltrating.
These results predict that compressed air will movetowards a drained boundary and be finally exhausted.
However, further considerations are needed to clarify
the precursory phenomena of landslides.
REFERENCES[1] Tsuchida, T., Yuri, A., Kano, S., Nakayabu, K.,
Yabuki, K., Hanaoka, N. & Kawabata, S.
‘Consideration of odor intensity within the
ground and odor generation on landslide’,
Journal of JGS, Vol.8, No.2, 2013, pp.339-348.
[2] Kanazawa, S. ‘Numerical evaluation of
mechanical behavior of compacted soil
structures based on a mathematical modelling forunsaturated soils’ Doctoral dissertation,
Department of Civil Engineering, Kobe
University, 2010 (in Japanese).
[3] Ohno S., Kawai K. & Tachibana S., ‘Elastoplastic consti-tutive model for unsaturated soil
applied effective degree of saturation as aparameter expressing stiffness’ Journal of JSCE,
Vol. 63, No.4, 2007, pp.1132-1141 (in Japanese).
[4] Borja, R. I., ‘A mathematical framework forthree-phase deformation and strain localization
analysis of partially saturated porous media’Computer methods in applied mechanics andengineering, Vol.193, 2004, pp.5301-5338.
[5] Mualem, Y., ‘A new model for predicting thehydraulic conductivity of unsaturated porous
media’ Water Resources Research, Vol.12, No.3,
1976, pp.514-522.
[6] Van Genuchten, ‘A closed-form equation forpredicting hydraulic of unsaturated soils’ Soil
Science Society American Journal, Vol.44, 1980,
pp.892-898.
[7] Kawai, K., Iizuka, A., Hayakawa, E. & Wang, W.
‘Non-uniform settlement of compacted earth
structures caused by the deformation
characteristics of unsaturated soil on wetting’Soils and Foundations, Vol.47, No.2, 2007, pp.
195-206.

238

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

CYCLIC LATERAL RESPONSE OF MODEL PILE GROUPS FOR
WIND TURBINES IN CLAY SOIL
Werasak Raongjant 1, Meng Jing 2
1

Assistant Prof., Department of Civil Engineering, Faculty of Engineering,
Rajamangala University of Technology Thanyaburi, Thailand.

2

Assistant Prof., Department of Civil Engineering, Faculty of Engineering,
Rajamangala University of Technology Thanyaburi, Thailand.

ABSTRACT：Due to the large diameter and the cyclic loading, it is necessary to investigate the lateral
response of pile groups for wind turbines in clay soil. This paper presents an experimental research on 2x2
model pile groups embedded in different types of clay soil, relatively density of 90% and 10%, subjected to
one-way lateral cyclic loading. Analysis on the experimental results, such as load–deflection curves andbending moment profiles, were processed. It is found that the magnitude of the head level deflections and the
maximum bending moment increased with the number of loading cycles. The group interaction effect undercyclic lateral loading is more significant in loose soil.
Keywords: Pile groups, Cyclic Lateral Loading, Relatively Density, Wind Turbines, Model Test

Limit research experiences of pile group
foundations under cyclic lateral loading could be
reviewed. Rollins et al. [4] tested series of fullscale pile groups with various spacing to study theeffect of pile spacing under cyclic lateral loadingon the behavior of pile groups. It was concluded
that the group interaction effect decreases
considerably with increase of spacing from 3.3 to
5.65 times the diameter of pile. Peng et al. [5]summarized various devices used for applying
cyclic lateral load to model piles. Chandrasekaran
et al. [6] reported experimental investigation on
1×2, 2×2 and 3×3 model pile groups, having
different length or spacing and embedded in soft
marine clay under lateral cyclic loading. From thetest results it could be seen that, the lateral capacityof the pile group increased with the number of pile.
Piles in closely spaced groups carry considerably
higher moments at deeper depths. Basak S. and
Mastorakis Nikose [7] performed the experimenton 2×2 pile group, each pile being hollow circular
stainless steel bar. During cyclic loading inprogress, a pair of gaps appeared in front and theback of the pile with the increasing number ofloading cycles. The degradation factor decreased
with number of cycles and increased withfrequency nonlinearly.
Also there are some researches on the lateralbehavior of pile groups using numerical analysis.
Abbas et al. [8] reported the numerical studies on
the lateral behavior of three pile groupconfigurations (i.e. 2×1, 2×2 and 3×2 pile groups)
with four values of pile spacing (i.e. 2D, 4D, 6D

1. INTRODUCTIONWind energy is now the popular energy
resource all over the world due to its clean andrenewable advantages. Also in Thailand, thegovernment has planned to install the wind energy
capacity to 800 MW by the year 2022 [1]. The pilegroup foundation consisting of a number ofsmaller piles connected by a platform plate is oneof the suitable solutions for supporting windturbine. The pile groups supporting thesestructures are subjected to cyclic lateral loading
generated by wind, which is quite large and will
governs the design of pile groups. Pile group
behavior under cyclic lateral loading is nonlinear
and involves complicated group interaction [2],[3].
Inevitably, group interaction has significant effecton the behavior of pile groups also.

Fig. 1 The Danish offshore wind farm in the North
Sea [1]
239

GEOMATE- Bangkok, Nov. 14-16, 2016

𝑬𝑬𝒎𝒎 𝑰𝑰𝒎𝒎
𝟏𝟏= 𝟓𝟓
𝑬𝑬𝒑𝒑 𝑰𝑰𝒑𝒑𝒏𝒏
in which E m is the modulus of elasticity of model
pile, which was recorded as 69 GPa for aluminumin this study; E p is the modulus of elasticity of
prototype pile, corresponding to the reinforced
concrete of f c ’ = 25 N/mm2 in this study; I m is themoment of inertia of model pile; I p is the momentof inertia of prototype pile; and 1/n is the scale
factor for length of model pile to prototype pile,which equal to 1/30 in this study.

and 8D). It can be observed that, the groupconfiguration has significant effect on the lateralpile displacement and ultimate soil resistanceunder the same lateral load. Chore et al. [9] usedfinite element method to analyze two pile groups
subjected to lateral loads by considering the nonlinear behavior of soil. It was concluded that the ofsoil due to non-linear behavior of soil, the top
displacement of the pile group was increased
66.4% to 145.6%, while the fixed moments was
reduced 2% to 20% and the positive moments 54%
to 57%.
There are limited literature presented theexperimental results of the pile groups in clay soil
with different relative density under cyclic lateralloads. The effects of soil on pile groups undercyclic loading have not previously been studied
extensively. Hence, in the present study, fourmodel pile groups embedded in clay soils were
tested under one-way cyclic lateral load to study
the effects of soil density and number of cycles on
load–deflection and bending behavior of pile group.
2. TEST PROGRAM
2.1 Soil Used
Clay used in this study is mined from the area
in Pathum Thani province in Thailand. The
properties of the clay are: liquid limit = 54.5%,
plastic limit= 35.48%, plasticity index= 18.92%.
According to the determined value, dense soil withrelative density of 90% and loose soil with relative
density of 10% were prepared in the test, following
the compaction standard method. Clay was put
into the test tank and compacted well. The
placement density of dense clay soil (relative
density of 90%) is 1.425 g/cm3. The placement
density of loose clay soil (relative density of 10%)
is 1.253 g/cm3. The conformed moisture content ofboth type of soil is 29.5%, according to that in
Pathum Thani province of Thailand.

Fig.2 Model pile
750 mm long aluminum stick having squaresection of 10×10 mm was used as model pile toarrange 2×2 pile group with spacing of 63 mm inthis study. Aluminum plate in 93× 93cm squaresection and in 15 mm thick was used as pile cap.
Aluminum stick in size of 30×30 mm and 150 mm
long was welded above the pile cap to subject
lateral load from hydraulic jack. Four strain gauges
at the interval of 150 mm were pasted on selected
pile in leading row or rear row to record thebending moment at different depth of pile groups(see in Fig.2).

2.2 Pile Groups

2.3 Test Loading Pattern

The origin of the model pile group is the pilegroup foundation for wind turbine with the towerof 25 m high and the rotor diameter of 10 m. Fourconcrete piles, 19 m long and in 50×50 cm solid
section, were arranged in the case of 2×2 pilegroup with the hanging space of 1.9 m. The pile
cap is in 2.80×2.80 m square section with the
thickness of 45 cm. Similitude laws proposed by
Wood DM and Crewe A. [10] are followed to
select the material and dimensions of model pile.

The steel tank used in test is 750 mm high withthe section diameter of 650 mm (see in Fig.3). The
model pile was then plug into the soil, leaving 120
mm above the soil surface. Then the cyclic lateralload was then applied by pneumatic cylinder
attached on the loading frame. The one-way cyclic
lateral load was increased from a minimum valueof P min = 0 to a maximum of P max =20N, andreturn to P min = 0 (no reversal of loading direction).
The maximum lateral load P max was taken as 80%of the ultimate lateral loading capacity of 25N,
which has determined from static test when the240GEOMATE- Bangkok, Nov. 14-16, 2016

corresponding deflection is equal to 20% of pile
diameter. The lateral load was repeated for 50
times (cycles) and then increased monotonically
until failure.

deflection increased nonlinearly with the numberof loading cycles. It shows that, for pile groupembedded in dense soil, the load-deflection curves
under cyclic or static load behaved similar trend.
The deflection of pile group in dense soil after 50
loading cycles was about 4.27 times more than thevalue for the same static load. However, thedeflection of pile group in loose soil after 50
loading cycles was about 65% of the value for thesame static load. It can also be seen that the
deflection of pile group in dense soil, developed in
the first 15 loading cycles, is accounted for 66% of
the deflection at 50 cycles. For pile group in loose
soil, this value is 50%.
3.2 Bending Moment Profiles
Based on the strain data measured along thedepth of the pile in leading row and rear row of thepile group, the bending moment at various depth
can be calculated by the expression: 𝑀𝑀 =𝐸𝐸𝐸𝐸𝐸𝐸𝑟𝑟, inwhich E is the young’s modulus of the model pile;
I is the moment of the inertia of the model pile; Ɛis the measured strain; and r is the horizontal
distance between strain gauge position and neutral
axis, equal to half width of the section of pile.

Fig.3 Test set-up
3. DISCUSSION ON TEST RESULTS

3.2.1 Effect of loading cycles
3.1 Load-Deflection Curves
60
50

0

Bending moment (N-cm)
20
40

60

0
10

cyclic2040Depth (cm)Load (N)-20loose soil

30
20
static103040cycles 1
cycles 10
cycles 20
cycles 30
cycles 40
cycles 50

50
60

0

0

10

20

70

30

Deflection (mm)80(a) Loose soil
Dense soil
60

-20
staticBending moment (N-cm)
20
40

60

80

0

40

10

30

20

cyclic20

Depth (cm)Load (N)

50

0

10
0
0

10
20
Deflection (mm)

30

30
40
50
60

Fig. 4 Load versus head level deflection curves ofpile groups in different soil

70

cycles 1
cycles 10
cycles 20
cycles 30
cycles 40
cycles 50

80

The lateral deflection of model pile groups
under lateral static or cyclic loading, measured at
the level of pile caps, were drawn in Fig. 4. The(b) Dense soil
Fig.5 Bending moment of piles in dense or loose
soil
241

GEOMATE- Bangkok, Nov. 14-16, 2016

For the cyclic loading test, Fig.5 presents lateral
response of model pile at the same position in pilegroups. The bending moments at various depths
increased with the number of loading cycles. From
50cm below the soil surface down, the pile indense soil subjected to bending moment in inverse
direction. The maximum bending moment in pileembedded in loose soil after 50 loading cycles was
about 1.6 times of that at the first cycle. However,for pile in dense soil, the maximum bending
moment for 50 cycles was 2 times of that at thefirst cycle. The depth corresponding to the
maximum bending moment acted on pile group inloose soil at 50 loading cycles were measured atabout 5 cm below the corresponding depth at thefirst cycle. However, marginal increase in thedepth corresponding to the maximum bending
moment was observed for pile embedded in dense
soil.

cycles. It is noted that, for pie in leading row of thepile group, the maximum bending moment
subjected by pile at 50 loading cycles in either
loose or dense soil was more than 2 times of thevalue for the same static load. Moreover, for pilein loose soil, the maximum bending moment
occurred at the deeper depth. For pile in rear row
of the pile group, the maximum bending moment
subjected by pile at 50 loading cycles in either
loose or dense soil was more than 6 times of thevalue for the same static load. However, depth ofthe maximum bending moment undergoes littlechange.
3.2.3 Effect of the position of pile in pile group
Bending moment (N-cm)
-20

0

20

40

60

80

0
10

3.2.2 Effect of loading program-20

0

Bending moment (N-cm)
20
40
60

Depth (cm)

20
80

0

Depth (cm)

10

30
40

20

60

30

70

40

80

50

Loose soil

Loose static
Dense static
Loose cyclic
Dense cyclic

60
70

-20

80

0

Bending moment (N-cm)
20
40
60

80

10

30
40
50

Depth (cm)

20
60
30
70

40

70

80

20
Depth (cm)

0

60

60

10

0

50

Bending moment (N-cm)
20
40

0

(a) Pile in leading row of the pile group-20

pile1 cycle=20
pile2 cycle=20
pile1 cycle=50
pile2 cycle=50

50

pile1 cycle=20
pile2 cycle=20
pile1 cycle=50
pile2 cycle=50

80

Dense soil
Fig.7 Bending moment for piles in leading or rearrows

Loose static
Dense static
Loose cyclic
Dense cyclic

80

The bending moment profiles for piles atdifferent positions in the pile groups embedded indense or loose soil are shown in Fig.7. Morebending moment was carried by pile 2 in rear row
under cyclic lateral loading. When embedded inloose soil, the maximum bending moment
subjected by pile 2 at 20 loading cycles was 50%
more than that of pile 1 in leading row. The(b) Pile in rear row of the pile group
Fig.6 Bending moment behavior for pile groups
under static and cyclic loading (cycles = 50)
Fig.6 shows the bending moment along thedepth of piles under static and cyclic loading at 50
242

GEOMATE- Bangkok, Nov. 14-16, 2016

magnitude of this difference decreased to 5% at 50
cycles. When embedded in dense soil, themaximum bending moment subjected by pile 2 at
20 loading cycles was 38% more than that of pile 1
in leading row. The magnitude of this differencechanged little at 50 cycles. This indicated that
more load was reacted by the rear row piles undercyclic lateral loading. Especially for pile groupembedded in dense soil, the design will begoverned by the pile in the rear row.

4. CONCLUSION
Based on the discussion of the test results
aforementioned,someconclusions
weresummarized here.
(1) The deflection of pile group in dense soil after
50 loading cycles was about 4.27 times more thanthe value for the same static load. However, thedeflection of pile group in loose soil after 50
loading cycles was about 65% of the value for thesame static load.
(2) With the increasing of loading cycles, thedeflection of pile groups increased quickly to acertain value. Then after 22 cycles, the increasing
became slow. More than 50% of the total
deflection at 50 cycles occurred in the first 15
cycles.
(3) More loads were reacted by the rear row piles
under cyclic lateral loading. For pile in rear row of
the pile group, the maximum bending moment
subjected by pile at 50 loading cycles in either
loose or dense soil was more than 6 times of thevalue for the same static load. Especially for pilegroup embedded in dense soil, the design will begoverned by the pile in the rear row.
(4) The depth corresponding to the maximum
bending moment acted on pile group in loose soil
at 50 loading cycles is deeper than that at the firstcycle. However, marginal increase in the depth
corresponding to the maximum bending moment
was observed for pile embedded in dense soil.

3. 3 Effect of soil
Development of the deflection with the numberof loading cycles for pile groups in dense or loose
soil is shown in Fig.8. The deflection increased
nonlinearly with the number of loading cycles. It
can be seen that, with the increasing of loading
cycles, deflection of pile groups increased quickly
to a certain value. Then after 22 cycles, theincreasing became slow. It can also be observedthat the soil had a significant effect on thedeflections. As the relative density of soil decrease
from 90% to 10%, there was approximately 4% to
16% increase in the deflection level. The
deflection for pile group in loose soil is 6.16 mm
and 15.96 mm for the first and 50th cycles,
respectively. The corresponding values for dense
soil were 3.42 mm and 13.60 mm, respectively.
Plastic soil deformation around the pile was noted(Fig.9). A ‘gap’ occurred under the cyclic loading.

5. ACKNOWLEDGEMENTS

18

Deflection (mm)16

The work described in this paper was carriedout with generous financial assistance RMUTTAnnual Government Statement of Expenditure in2014 provided by the Rajamangala University ofTechnology Thanyaburi (Thailand).

14
12
10
8
6
Loose
Dense

4
2

6. REFERENCES

0
0

10

20
30
Number of cycles

40[1] Das Manish Kumar and Leephakpreeda
Thananchai, “A review study on trends to windenergy in a global and Thailand context”, Sura. J.
Sci. Tech. 2011 (1), pp.1-13.
[2] Cuellar Pablo, “Pile Foundations for Offshore
Wind Turbines: Numerical and Experimental
Investigations on the Behaviour under Short-termand Long-term Cyclic”. Doctor Thesis, 2011,
Berlin University of Technology, Germany.
[3] Li Z., Haigh S. K. and Bolton M. D.,“Centrifuge modeling of mono-pile under cyclic
lateral loads”. In: 7th International Conference on
Physical Modeling in Geotechnics, Zurich. 2010.
Vol. 2, pp.965-970.
[4] Kyle M. Rollins, Ryan J. Olsen, Jeffery J.
Egbert, “Derek H. Jensen, Kimball G. Olsen and
Brian H. Garrett. Pile spacing effects on lateral pile50

Fig.8 Deflection for pile groups in different soils

Fig.9 Gap occurred under cyclic loading243

GEOMATE- Bangkok, Nov. 14-16, 2016group behavior: load tests”. J. Geotech.
Geoenviron. ASCE (2006a), 132(10), pp.1262–
1271.
[5] Peng J.R., Clarke B.G., Rouainia M., “A
device to cyclic lateral loaded model piles”.
Geotech. Test. J., ASTM. 2006, 29(4), pp.1-7.
[6] Chandrasekaran S., Boominathan A. andDodagoudar G. R., “Experimental investigations
on the behaviour of pile groups in clay underlateral cyclic loading”. Geotech. Geol. En. 2010,
Vol. 28, pp.603–617.
[7] Basak S. and Mastorakis Nikose,“Experimental investigation of pile group underlateral cyclic load in soft cohesive soil”.
Proceedings of the 4th IASME/WSEAS
International Conference on Continuum Mechanics(CM'09). 2009. pp.50-55.

[8] Abbas Jasim M, Chik Zamri and Tah Mohd
Raihan, “ Influence of group configuration on thelateral pile group response subjected to lateralload”. Electoron. J. Geotech. En. 2010, Vol. 15,
pp.761-772.
[9] Chore H.S., Ingle R.K. and Sawant V.A.,“Non-linear analysis of pile groups subjected tolateral loads using ‘p-y’ curve”. Inter. Multi. Mech.
2012, Vol. 5, pp.57-73.
[10] Wood DM, Crewe A., “Taylor C. Shaking
table testing of geotechnical models”. Int. J. Phys.
Model. Geotech.. 2002, Vol 1, pp.1-13.

244

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THE USE OF POLYURETHANE TO MAINTAIN STRENGTH OF
ROAD SUBGRADE FOR FLOOD DAMAGE CONTROL
Safawati Mohd Radzi1, Abdul Naser Abdul Ghani2, Muhd Shahril Nizam Ismail1, Ahmad Hilmy Abdul
Hamid2 and Khalid Ahmad2
1
Postgraduate, School of Housing Building and Planning, Universiti Sains Malaysia, MALAYSIA
2School of Housing Building and Planning, Universiti Sains Malaysia, MALAYSIA

ABSTRACTThe strength of subgrade soil or foundation can influence the design of road pavement structures. Flood canbe one of the causes of road damages. Since the condition of subgrade layer is critical in the road pavement
stability, a study was carried out to ascertain the use of polyurethane insertion as a stabilization mechanism inroad subgrade. This study was conducted based on two types of soil that are usually used as soil embankment inroad construction. California Bearing Ratio (CBR) test was conducted to the various categories of soaking days
and repeated submerged conditions to determine the strength of subgrade soil with and without polyurethane
layer. It can be concluded that polyurethane layer can be used to increase or maintain the strength of subgrade
soil from the inundation effect.
Keywords: Subgrade, Road, Flooding, California Bearing Ratio (CBR)

stabilization is one method for soil improvement
which help to improve the subgrade soil strength
hence reducing the impact of flooding on roadinfrastructures. There are few types of existing
subgrade improvement using chemical, geotextile
and mechanical. Poor quality of the subgrade soil
can be improved by the treatment of the soil using
stabilizing agents cement or lime [5], [6]. For
example, soft clay mixed with cement will
strengthened the subgrade because cement and water
react to form cementitious calcium silicate andaluminum hydrates which can bind the soil particles
together [7].
Geotextiles can also be used for subgrade
stabilization because it can maintain physically the
integrity of pavement layer boundaries and it enables
water to move in an unrestricted manner. Subgrade
alone, when it is weak, cannot support layerinterface because of the interpenetration of soft
subgrade with granular layer under the high trafficloading. As solution, geotextiles can be placed at aninterface of the soft subgrade and granular layer toprevent the two materials intermixing and function
to stabilize the interface between two materials andenable the granular layer to maintain the compacted
strength [8].
There are many application of polyurethane in
the manufacturing and construction industry.
Polyurethane has been used to improve the safety
and performance of railway track infrastructure dueto the stiffness and ductility [9] and has become asolution to the settlement problem with the injectionof polyurethane to the pile and slab in order toreplace the conventional underpinning pile [10].

INTRODUCTION
Floods have great impacts on road infrastructure
and people as their activity may be disrupted and theimpacts, in most cases, can last for more than one
week. In the coming years, climate change could
make the situation become more challenging [1].
Floods are also among the most costly natural
disasters in terms of returning occurrence, economic
loss and human suffering. These flooding damage
the roads, environmental surrounding, and livelihood.
The impact of the flood can take the long term effectin maintenance works since the whole range of civil
infrastructures are usually involved. The continuous
flood submersion to the roads would bring damage
on large part of any city or village infrastructures [2].
However, damage to road pavement and roadway
could create disconnection of people movement andemergency supply or evacuation works. Road
subgrade is the most prone to the flood effectbecause it is at the lower level of road structure,having the largest exposure to flood. Ghani et al. [3]found that the CBR values for subgrade strength
decreases due to the higher number of inundation
days of the subgrade soil.
Subgrade soil can be strengthened by thestabilization agents such as cement, lime and fly ash.
Besides the three agents, polymer or polyurethane
has also been used to stabilize sand clay mixtures inthe soil slope erosion [4]. The impact of flooding
can be disastrous to the infrastructure and theenvironment. On top of that there will be huge
expenditure for the rehabilitation process and it may
take some time to make things back to normal. Soil
245

GEOMATE- Bangkok, Nov. 14-16, 2016

Polyurethane has also been used to fill void in thesoil and reduced settlement issues [11]. The
advantage of polyurethane is it is durable and haveshorter time to complete remediation which may
gain 90% compressive strength within 15 minutes
from injection. In order to determine the feasibilityof using polyurethane in road subgrade, there is aneed study on the existing application ofpolyurethane related to soil and how it can beapplied in case of road subgrade.
MATERIALS AND METHOD
The use of polyurethane is simulated in the lab
using CBR mould and test. Two (2) types ofcommonly used or found soil as subgrade layer wereused in the CBR test. Details of the soils are shownin Fig.1, 2, 3 and 4.

Fig 4: Soil 2
Other basic properties of the soils used in thisstudy are shown in the following Table 1.
Table 1: Basic Properties of Soils
Properties

Fig 1: PSD for Soil 1

Soil 1

Soil 2

Specific
Gravity

2.64

2.60

Maximum Dry
Density(kg/m3)

1964

1510

Plasticity
Index

Non Plastic

28.5%

Grading

Well Graded
Sand (SW)

Poorly Graded
Sand
With
Clay (SP-SC)

The Polyurethane
The polyurethane layer used in this study wasproduced from mixing two (2) compounds namely
polyol and polyisocyanates at a ratio of 1:1.2. The
compound that can be obtain from local supplier isshown in Fig 5 below.

Fig 2: PSD for Soil 2

Fig 3: Soil 1
246

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig 5: Mix ratio of polyurethane compounds
Cured polyurethane in a PVC mould were thencut into sizes as shown in Fig 6 for placement in the
CBR mould together with the subgrade soil samples
as shown in Fig 7.

Fig 8: Specimen under water to simulate flood
Specimens were kept under water according to
the specified inundation period of 1, 3, and 7 day
duration. Another set were repeatedly inundated forone hour only on day 1, 3, and 7 to simulate repeatedflood submerge.
California Bearing Ratio (CBR) tests wereconducted on all the samples of soil according to BS
1377 using instrumented CBR equipments.

Fig 6: Polyurethane

RESULT AND DISCUSSION
Soil 1
Fig 9 shows the difference between the strengthof the soil that unsoaked and soaked conditions forsample of soil that normal and there is an additionalof polyurethane. The samples of soil left to soak inwater for different days which are 1, 3 and 7 days.
The graph shows the CBR value of the sample ofsoil that does not soaked is higher than the samples
of soil that soaked. This is because the saturated
period of the soil sample. The CBR value to thesample of soil that not immersed is 50.35%. The
result CBR values for the samples of soil that soakedfor 1, 3, and 7 days is 3.65%, 6.45% and 3.8%
respectively. These results show that the presence ofwater during the immersion causes the degradationof the soil sample.

Fig 7: Placement of Polyurethane Layer

Specimen and Test Preparation
CBR specimens with and without polyurethane
were prepared in the lab. Altogether 24 specimens
were prepared from Soil 1 and Soil 2 (12 each). The
specimens were intended for control, continuous
soaking and repeated submerge conditions. Fig 8shows specimens kept under water.

Fig 9: Inundation days and CBR strength (Soil 1)

247

GEOMATE- Bangkok, Nov. 14-16, 2016

strength of soil tends to decrease due to theinundation of the days.

However, with the presence of polyurethane in
the soil sample, the CBR value can be increasedslightly. This is means the strength of soil can beincreased. For the sample that not immersed in thewater, the CBR value is 72.27%. It shows asignificant increase to the soil sample. Besides that,in the 1 day of soaked condition, the value of CBRrise from 3.65% to 5.75%. On the other, CBR values
for soil sample 3 days of soaked is 6.75% and for 7
days of soaked is 6%. The CBR values for threesamples with polyurethane in the soaked conditions
have shown higher than the CBR values for normal
samples. It can be concluded that polyurethane canhelp to maintain or increase the strength of the soil.
Graph bar in Fig 10 shows the comparison on thesoil samples containing polyurethane between two
different situations which is repeated soaked andunsoaked conditions. The soil samples are soakedfor 1 hour at Day 1, Day 3 and Day 7.

Fig 1: Inundation days and CBR (Soil 2)
However, for the soil sample that containing
polyurethane, CBR value for unsoaked condition
seen as 36.06% which is obviously higher than CBRvalue for soaked conditions. In the graph, the CBRvalue for 1 day soaked is 1.52%, while CBR value
for 3 days soaked is 1.89% and for 7 days soaked,
the reading of CBR is 1.74%. Generally, the value of
CBR for soil samples in 3 and 7 days soaking period
is increase.
Subsequently, in Fig 12, the graph shows theresult of Soil 2 samples for CBR value for control
and repeated submerged sample containing
polyurethane. The repeated submerged sample issoaked for 1 hour on Day 1, Day 3 and Day 7. The
control sample shows the highest result with CBRvalue of 36.06% compare to the repeated submerge
sample. CBR value for the repeated submerge
samples are decreasing respectively from day 1 today 7 in which the result are 2.35% on Day 1. Then
it is reduced to 1.65% on Day 3 and 1.30% on Day 7.
So, it can be seen that CBR value become lower
when the soil was submerged again and again underwater.

Fig 10: Repeated submergence CBR (Soil1)
The CBR values on the bar chart show thedifferent patterns of the result. It shows that theresult on unsoaked condition is 72.27% and for the
repeated submerged condition for 1 hour on Day 1,
Day 3 and Day 7, the CBR result is 8.35%, 5.1% and6.95% respectively. In this repeated submerged case,
the CBR value was reduced on Day 1 and the value
subsequently also reduced on Day 3 compared to theunsoaked sample soil. It can be seen that the CBRvalues are compatible with the continuous
inundation specimens from Fig 9 before.
Soil 2
Fig 11 shoes the comparison of the CBR value
for the soil sample that contains polyurethane andthe normal based on the two different conditions
which are for unsoaked and soaked condition. For
the normal sample soil, the bar graph shows the
CBR value for unsoaked condition higher than the
CBR value to the soaked conditions. The strengthfor unsoaked condition soil sample is 44.39% while
the CBR value for the soaked condition for 1, 3 and7 days is 4.3%, 1.7% and 1.3% respectively. The

Fig 12: Repeated submergence CBR (Soil 2)
Other Observation
It is also important to note that the movement of
the CBR plunger downward is also contributed by
the compressible effect of the polyurethane layer
itself. This can be observed when the soil and the248

GEOMATE- Bangkok, Nov. 14-16, 2016

polyurethane layer are removed from the mould asshown in Fig. 13 and 14. If this movement can bepredicted and control, the ultimate actual subgrade
strength improvement is actually higher than what isshown above.

ACKNOWLEDGEMENTS
This work is supported by the Universiti Sains
Malaysia
RUI
Grant
SchemeNo.
1001/PPBGN/814283.

REFERENCES[1] NFRA, Flood National Risk Assessment.
December 2011.
[2] Bowoputro, H. et al. (2009): Temperature
Influence and Lapindo Mud Submersion
towards the Stability Value of Asphaltic
Concrete Mix, Rekayasa Sipil Journal.
[3] Ghani, A.N.A., Roslan, N.I.,and Hamid, A.H.A.,(2016). Road Submergence During Flooding
and Its Effect on Subgrade Strength. Int. J. of
GEOMATE, May 2016, Vol.10, Issue 21, pp
1848-1853.
[4] Liu J, Shi B, Huang H, Jiang H (2009)
Improvement Of Water-Stability Of Clay
Aggregates Admixed With Aqueous Polymer
Soil Stabilizers.
[5] Bell FG (1995) Cement Stabilization and Clay
Soils, with Examples. Environmental &Engineering Geoscience.
[6] Bell, F.G. (2007). Engineering Geology. 2nd
Edition, Elsevier Science Publishers, Oxford,
United Kingdom[7] Gueddouda, M. K., Goual, I., Lamara, M.,
Smaida, A., & Mekarta, B. (2011). Chemical
Stabilization of Expansive Clays from Algeria,
Journal Of Researches In Engineering. 11(5).
[8] Lawson, C.R., (1995), “Subgrade Stabilisation
With Geotextiles”, Geosynthetics International.
[9] Woodward, P.K., Kennedy, J. And Medero, G.
(2009).
Three-Dimensional
Track
Reinforcement Using Polymer Geocomposites,
Proceedings Of The 2009 American Railway
Engineering And Maintenance Of Way
Association (AREMA), Chicago, Illinios, USA[10] Philip. M. (2010). Uretek Ground Engineering
http://www.uretek.com.au/pages/pdf/32_Page_
Book.pdf[11] Diana, Ismacahyadi, Nor Azizah, Mohamed
Khatif (2014) Evaluation Of Performance For
Polyurethane Injection Pile System As A
Ground Improvement Using Plaxis 2nd
International Conference On Innovation And
Technology For Sustainable Built Environment.

Fig 13: Polyurethane layer before test

Fig 14: Polyurethane layer after test becomes thinner

CONCLUSION
This experimental study had been conducted inorder to determine the strength of soil sample
containing polyurethane and its capability tomaintain the strength when the soil samples are kept
in different inundation conditions. From the result of
this study, it can be concluded that the soil samples
that have been kept in longer duration of inundation
tends to loss the strength as expected. Polyurethanethat was placed in the soil samples and tested in the
different days of inundation indicated that it can help
to increase and maintain the strength of the soil. The
reduction of subgrade strength once it is inundated
can be very significant. In this study, polyurethane
layer helps recover at least 3% of the strength loss
due to inundation.
However, there is one issue that requires further
investigation. The issue is about the compressibilityof the polyurethane layer under load. This issue must
be comprehended and resolved before actual
application on site.

249

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EVALUATION OF CONSTRAIN MODULUS-BULK STRESSRELATIONS OF PAVEMENT STRUCTURE MATERIALS BY USINGCBR MOULD
Sawanya Dararat1 and Warat Kongkitkul2
Department of Civil Engineering, Faculty of Engineering,
King Mongkut’s University of Technology Thonburi, Thailand
1,2

ABSTRACT
Resilient modulus (MR) of pavement structure materials is an important parameter used in the mechanistic
pavement engineering design. In general, MR is evaluated by applying a number of repeated loads at various stressstates in a triaxial test which is complicated and expensive. This paper presents an alternative method to determine
the constrain modulus (M) of a dry sand, a lateritic soil and a crushed rock, under one-dimensional deformation.
At various vertical stress levels, M values were determined by applying small strain-amplitude cyclic loadings to
the compacted soil specimen prepared in a CBR mould. In the present study, the CBR mould was made special inthat it can measure lateral stress confined the specimen during a test. Hence the bulk stress () can then be
determined, and M- relations for the tested materials were presented. In addition, an analytical method for
eliminating the effects of bedding error was attempted so as to obtain the true M value. It is found that M is notconstant but increases with , similar to MR- relations found with the resilient modulus test. And bedding error
is important and can result in significant underestimation of the true M value.
Keywords: Pavement structure materials, Constrain modulus, Resilient modulus, Coefficient of lateral earthpressure, Bedding error

directly from the specimen prepared with the standard
CBR mould.
To this end, many researchers have developed
alternative methods to evaluate the material’s
stiffness for the pavement design purpose, forexample, the K-Mould [8], Springbox [9], and PUMA[10]. One-dimensional cyclic loading test on a
specimen prepared in a CBR mould is easy to conduct
in laboratory. Then constrain modulus (M) can be
determined. However, this test method could notmeasure the lateral stress. It is therefore necessary tomodify the standard CBR mould so that themeasurement of lateral stress is possible. In this study,
a CBR mould was attached with strain gauges tomeasure the hoop strain and converts to the lateralstress. Then the coefficient of lateral earth pressure atrest (k0) was determined. Hence, the bulk stress (),which is summation of all the normal stresses, canthen be determined. Then, at the same , M and MR
could be compared, and if there are correlations, MR
could be estimated from M which can be reliably
determined in a much easier method. In addition, theusual measurement of specimen’s deformation
always includes the so-called bedding error whichsignificantly affects the determined material’s
stiffness. In this study, it was attempted to prepare test
specimens with different heights and used thecorrection procedures proposed by Koseki et al. [11]to determine the true M value that is free from any
bedding error.

INTRODUCTION
Resilient modulus (MR) of pavement structurematerials is an important parameter used in themechanistic pavement engineering design. Resilient
modulus is a property of the pavement systemcorresponding to repeated traffic loadings. The
resilient modulus laboratory testing of soil andaggregate materials has been recommended by
AASHTO T307 [1]. Triaxial apparatus is widely used
for evaluation of resilient modulus of pavement
structure materials [2]-[5]. To utilise the MR fromtriaxial results, k- model is the most commonly used
for granular materials in the pavement engineering
design (Eq.1).

M R =K1θ K2(1)where K1 and K2 are regression constants
However, preparation of test specimens andshearing loading history employed in the triaxial tests
for evaluation of MR are complicated, and therefore itis difficult, expensive and time consuming. On theother hand, the resilient modulus can be estimatedfrom correlations with California Bearing Ratio(CBR), as in [6], [7]. It is of great interest in thepresent study to determine the resilient modulus or asimilar parameter that reflects the elastic stiffness
250

GEOMATE- Bangkok, Nov. 14-16, 2016

Summarising the above, in this study, it was
attempted to evaluate the constrain modulus ofpavement structure materials, and eliminate thebedding error from one-dimensional cyclic load tests.
A special CBR mould was used to evaluate lateralstress confined to the test specimen.

the latter loading pattern, monotonic loading isapplied firstly until the target vertical stress hasreached, and then sustained loading is performed for
30 minutes. Next, cyclic loading with a double stressamplitude of 30 kPa are applied for 10 cycles,
subsequently monotonic loading is applied again to
the next target vertical stress (Fig 4). Test program in
the present study is shown in Table 2.

MATERIALS AND APPARATUSES
Test Materials

Table 1 Compaction lateritic soil and crushed rockThree materials used in the test program were
KMUTT sand, lateritic soil, and crushed rock. Fig 1
shows their particle size distributions. KMUTT sandis in air-dry condition and used to verify the test
program and the special CBR mould (explained later).
The lateritic soil and crushed rock materials areprepared at 100% of maximum dry density (MDD)
and optimum moisture content (OMC). The
compaction curves are shown in Fig.2. Table 1 lists
the OMC and MDD of lateritic soil and crushed rock.

Materials

OMC (%)

Lateritic soil
Crushed rock

7.22
6.26

MDD
(g/cm3)
2.163
2.289

100

KMUTT sand
(D50= 0.3 mm)

80

Percent passing

Apparatuses
A cylindrical metal mould with an inner diameterof 152.4 mm and a height of 177.8 mm (without thecollar) was used. This mould is in accordance ASTM
D 1883-99. By using this mould, the lateral
deformation of specimen is confined (1D condition).
A special CBR mould that was attached strain gauges
is shown in Fig.3. To measure the hoop strain, twostrain gauges attached on side of the mould, and thenconnected with other two fix resistors to form a full
Wheatstone bridge circuit (opposite side 2-activegauge) (Fig. 3).

Lateritic soil
(D50= 3 mm)

60

40

Crushed rock(D50= 4.4 mm)

20

0
0.1

1

10

Paticle size in millimeters

Fig. 1TEST METHOD

Particle size
materials.

distributions

Lateritic soil
3
MDD = 2.163 g/cm
OMContent = 7.22 %Crushed rock3
MDD = 2.289 g/cm
OMC = 6.26 %oftested2.45
2.40

Specimen Preparation
Dry density , g/cm

3

2.35

A KMUTT sand specimen was prepared bypluviation through air. The density thus obtained isaround 1.52 g/cm3. The lateritic soil and crushed rockmaterials were prepared by compacting thesematerials that were laid layer-to-layer into the mould.
Three heights of specimens of 177.8 mm, 142.2 mm,
and 106.7 mm were prepared.

2.30
2.25
Crushed rock

2.20
2.15
2.10

Lateritic soil

2.05
2.00
1.95
2

4

6

8

10

Moisture content, W (%)

Test Program
Fig. 2

Two loading patterns were employed as shown in
Fig. 4. They are: i) continuous monotonic loadingwith a constant strain rate (ML); and ii) sustained
loading and then followed by cyclic loading (CL). In

251

Compaction curves of lateritic soil andcrushed rock.

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 Test program used in the present study
Materials

Initial height,
Load
H
Patterns, LP
KMUTT
H
ML
Sand
H, 0.8H, 0.6H ML, SL ,CL
Lateritic soil
HML
H, 0.8H, 0.6H ML, SL ,CL
Crushed rockH
ML
H, 0.8H, 0.6H ML, SL ,CL
Note: H = Initial height specimen is 177.8 mm

out

500

LOAD PATTERNS

450

in

Stage1, q at 50 kPa
Stage2, q at 100 kPa
Stage3, q at 150 kPa
Stage4, q at 200 kPa

400

Vertical stress, v (kPa)

a)

350

SL for 30 min
CL 10 times q = 30 kpa

300

End of loadingii) SL&CL

i) ML

250
Stage4200
Stage3150
Stage2

100
Stage150

Cyclic loading (CL)Creep (SL)Monotonic loading (ML)0Elapsed time, t (not to scale)Fig. 4

Loading histories.

TEST RESULTS AND DISCUSSIONS
Monotonic Loading Test Resultsb)

1

Figures 5(a), 5(b), and 5(c) show the variation ofvertical stress (  V ) and horizontal stress (  h ) withvertical strain (  V ) from ML tests on KMUTT sand,lateritic soil, and crushed rock, respectively. In thesetests, stress-strain curves exhibit continuous strainhardening behaviour. Here, it was assumed in thepresent study that the tangential strain (   ) mobilisedon surface of the CBR mould during a test is uniform
along the height and the periphery. The   value
measured by the strain gauges is therefore therepresentative of the entire CBR mould. Then, thehorizontal stress can be calculated from Eq. 2, basedon the theoretical of stress in cylindrical.

2

1

2

c)
Fig. 3 Et
h = 

Configurations of the special CBR mould:(a) a full Wheatstone bridge with two
active gauges and two fix resistors; (b)strain gauges attached on the side; and (c)connections of strain gauges and fix
resistorsr(2)

where E is Young’s modulus, t is a thickness, and r isa radius of CBR mould. It may be necessary tocalibrate the horizontal stress determined by thetechnique described above with other relevant
techniques that can directly measure the horizontal
stress so as to assure the measurement accuracy andprecision. However, it is not presently known to the252

GEOMATE- Bangkok, Nov. 14-16, 2016

authors how this calibration shall suitably be
performed.
Coefficient of lateral earth pressure at rest
( K 0   h /  V ) of materials are shown in Figs. 6a, 6b,
and 6c, for KMUTT sand, lateritic soil, and crushedrock, respectively. For a simulation in the pavement
engineering analysis, K0 of the base and subbase
courses are typically selected at 0.30-0.42 [12], [13].
However, K0 of lateritic soil and crushed rockmeasured from this test are around 0.4-0.5. Further
investigation on accuracy of the use of lateral stressmeasurement technique used in this study isnecessary to find out the slight discrepancies
mentioned above.

Constrain Modulus from CL Tests
Figures 7(a), 7(b), and 7(c) show the unloadingbraches Nos. 6-10 at the stress level of 100 kPa for
KMUTT sand, lateritic soil, and crushed rock,respectively. The stress-strain behaviour along thesebranches is highly linear only for a smaller range ofstress increment of the stress-strain loop. Lines werebest-fitted to the respective unloading branches
presented in Fig. 7 to obtain the constrain modulus(M). Then, the relationship between M and bulk stress( θ= V  2 h ) where  h =K 0 V for each stress levelwith different materials are plotted in Fig. 8. It can be
500
One-Dimensional Loading Test
Sample: KMUTT sand
400

500

Horizontal stress, h (kPa)

Vertical stress, v (kPa)Monotonic Loading Test

Vertical stress, v (kPa)400

One-Dimensional Loading Test
Sample: KMUTT sand

300
v

200
h

100

0

Monotonic Loading Test
300
1
200
k0=0.2950
100

0
0

50

-100
0.000

0.002

0.004

0.006

Vertical strain, v

a)

100

150

Horizontal stress, h (kPa)a)
500

One-Dimensional Loading Test
Sample: Lateritic soil

500
One-Dimensional Loading Test
Sample: Lateritic soil

400
Horizontal stress, h (kPa)

Vertical stress, v (kPa)Monotonic Loading Test
300
v

200

Vertical stress, v (kPa)

400

h

100

0

Monotonic Loading Test
300
1
200
k0=0.5065
100

0
0.000

0.005

0.010

0.015

0.020

0

Vertical strain, v

b)

50

100

150

200

Horizontal stress, h (kPa)

b)

500

600

One-Dimensional Loading Test
Sample: Crushed rock

300
v
h

100

Vertical stress, v (kPa)Monotonic Loading Test

200

500

Horizontal stress, h (kPa)

Vertical stress, v (kPa)

400

0One-Dimensional Loading Test
Sample: Crushed rock
Monotonic Loading Test

400
300
1
200
k0=0.4461
100
0

0.000

c)
Fig. 5

0.002

0.004

0.006

0.008

0.010

0.012

Vertical strain, v

-100
-50

c)

Vertical stress and horizontal stress – axial
strain relations from monotonic onedimensional tests on: (a) KMUTT sand;(b) Lateritic soil; and (c) crushed rock.

Fig. 6

253

0

50

100

150

200

250

Horizontal stress, h (kPa)

Relationship between vertical stress andhorizontal stress of: (a) KMUTT sand; (b)lateritic soil; and (c) crushed rock.

GEOMATE- Bangkok, Nov. 14-16, 2016

Evaluation of Effects of Bedding Error on
Constrained Modulus

clearly seen that the M value increases significantlywith an increase in the stress level [2], [4], [6]. Due tothe errors of axial deformation found with tests on
KMUTT sand and lateritic soil using the specimens
with the heights of 0.6H and 0.8H, the M value canbe confidently determined only for the stress level of
100 kPa. The data points at the higher stress levels forthese two materials were obtained by extrapolations
using the relation obtained for the specimen with theheight of H to pass through the only data point
measured at the stress level of 100 kPa, as shown in
Figs. 8a and 8b.

In order to evaluate effects of bedding error on themeasured constrain modulus, the analyticalprocedures proposed by Koseki et al. [11] was used.
According to Eq.3, a constrain modulus of a bedding
error layer (M1) and the constrain modulus of thenormal layer (M2) can be evaluated from relationships
between 1/M0 and 1/H [11].

300
One-Dimensional Loading Test
Sample: KMUTT sandLoops: Constrain
Modulus (kPa)
613595871363238137690913363410137373

Vertical stress, v (kPa)

One-Dimensional Loading Test
Sample: KMUTT sand, H
ML -> SL-> CL100

Constrain modulus, M (MPa)

120
2

R
0.98934
0.99080
0.99073
0.989140.98445

'vp=100 kPa

80
Mavg.= 136,196 kPa

200
ML -> SL-> CL
H
100

0.8H

0.6H

Generate
Generate

0.1
60
0.00340.51579

M0,0.8H= 

0.51579

M0,0.6H= 

0.51579

0.2

0.3

0.4

0.5

Bulk stress, (MPa)a)

0.0036

M0,H = 

Vertical strain, v

a)

1000
One-Dimensional Loading Test
Sample: Lateritic soil
Loops: Constrain
Modulus (kPa)
611967371147858116066911893410113304

Vertical stress, v (kPa)

One-Dimensional Loading Test
Sample: Lateritic soil, H
ML -> SL-> CL100

Constrain modulus, M (MPa)

120
2

R
0.97960
0.97504
0.97779
0.981090.97030

'vp=100 kPa

80
Mavg.= 116,552 kPa

ML -> SL-> CL
H
100
0.8H
0.6H

M0,H = 
M0,0.8H= 

Generate
Generate

M0,0.6H= 

0.24758

0.24758

0.24758

100.1

600.00840.00851

Bulk stress, (MPa)b)

0.0086

Vertical stress, v (kPa)100

One-Dimensional Loading Test
Sample: Crushed rock, H
ML -> SL-> CL

Loops: Constrain
Modulus (kPa)
6
44825
7
45635
8
43881
9
46375
10
45919

Constrain modulus, M (MPa)Vertical strain, v

b)

2

R
0.99774
0.99787
0.99630
0.99752
0.99721

'vp=100 kPa

80
Mavg.= 45,327 kPa

H

0.8H

0.6H

M0,H = 
M0,0.8H= 

10

0.1
0.00680.00690.00700.00710.00720.00730.0074c)Vertical strain, v

Fig. 8Fig. 7

ML -> SL-> CL100

0.93203

0.93203

M0,0.6H= 
60c)

One-Dimensional Loading Test
Sample: Crushed rockRelationships between vertical stress andvertical strain during unloading branches
to determine the constrain modulus atstress level 100 kPa on: (a) KMUTT sand;(b) lateritic soil; and (c) crushed rock.
254

0.932031

Bulk stress, (MPa)

Relationship between constrain modulus
and bulk stress on: (a) KMUTT sand; (b)lateritic soil; and (c) crushed rock.

GEOMATE- Bangkok, Nov. 14-16, 2016

1  1
1  D501=+
M 0  M1 M 2  H M 2(3)

The value of constrain modulus are summarised inTable 3. The M2 values are significantly larger than
the respective M0 values. On the other hand, theconstrain modulus of bedding error M1 values are
noticeably lower than M2.
The M2 values of crushed rock at the stress levelsof 150 and 200 kPa exhibit negative. It may beaffected from larger particles [11]. However, all theM2 values obtained from the present study should becompared/calibrated with the triaxial test results.
However, this work is beyond the scope of the paper.

where M0 is a nominal value of the constrain modulus
evaluated in conventional manner for whole
specimen height, H is a height of a specimen, and D50
is a particle mean diameter. In the present study, M0
is therefore the measured M value shown in Fig. 8,
which is dependent on the specimen’s height, and M2
is the constrain modulus of test material that is freefrom any bedding error and independent of thespecimen’s height.
Figures 9(a), 9(b), and 9(c) show relationships
between 1/M0 and 1/H for KMUTT sand, lateritic
soil, and crushed rock, respectively. It can be seenthat value of 1/M0 increased with an increase with thevalue of 1/H. Lines were best-fitted by assuming thatslopes for different stress levels are the same.

0.025

0.020

150
200
Lateritic
soil

50
100
150
200

Crushed
rock

50
100
150
200

M1
(MPa)

M2
(MPa)

57, 68,85
82, 98,136
101, 120,
160
117, 140,
172
55, 57,
83
65, 68,116
72, 75,
119
77, 81,
121
15, 18,2030, 37,45
37, 56,72
42, 64,
93

0.33

128

0.33

322

0.33

769

1/M0 (1/MPa)

0.015
2

1/M0=0.915*1/H+0.0031; R = 0.97021
2

1/M0=0.915*1/H+0.0013; R = 0.97850

0.010

2

0.005

0.006

0.008

0.010

0.012

0.014

1/H (1/mm)

a)
0.04

0.33

2500

2.02

200

2.04

500

2.04

1000

2.04

3333

One-Dimensional Loading Test
Sample: Lateritic soil
ML -> SL-> CL

1/M0 (1/MPa)

100

M0
(MPa)

1/M0=0.915*1/H+0.0078; R = 0.98872

1/M0=0.915*1/H+0.0004; R = 0.99184

333

1.32

-14

2

1/M0=1.468*1/H+0.0020; R = 0.84473
2

1/M0=1.468*1/H+0.0010; R = 0.85126
2
1/M0=1.468*1/H+0.0003; R = 0.85663

0.006

0.008

0.010

0.012

0.014

1/H (1/mm)

0.10

One-Dimensional Loading Test
Sample: Crushed rockML -> SL-> CL

Stress Level 50 kPa
Stress Level 100 kPa
Stress Level 150 kPa
Stress Level 200 kPa
2

1/M0=3.650*1/H+0.0298; R = 0.99979

0.06

0.04
2

1/M0=3.650*1/H+0.0003; R = 0.99979
2
1/M0=3.650*1/H-0.0071; R = 0.99061
2
1/M0=3.650*1/H-0.0102; R = 0.99991

0.02

0.00
0.006

c)
1.22

1/M0=1.468*1/H+0.0050; R = 0.86057

b)

34

1.20

2

0.02

0.08

1.16

Stress Level 50 kPa
Stress Level 100 kPa
Stress Level 150 kPa
Stress Level 200 kPa

0.00

1/M0 (1/MPa)

KMUTT
sand

Stress
Level
(kPa)
50

ML -> SL-> CL

Stress Level 50 kPa
Stress Level 100 kPa
Stress Level 150 kPa
Stress Level 200 kPa
2Table 3 Nominal constrained modulus and estimatedresults for bedding error layers and normal
layer
Testedmaterial

One-Dimensional Loading Test
Sample: KMUTT sand

0.008

0.010

0.012

0.014

1/H (1/mm)-98
Fig. 9255

Relationship between 1/M0 and 1/H : (a)KMUTT sand; (b) lateritic soil; and (c)crushed rock.

GEOMATE- Bangkok, Nov. 14-16, 2016[4] Yoshida N, Sugisako Y, Nakamura H, andHirotsu E, “Resilient modulus of hydraulicmechanically stabilized slag base-course
material”, Deformation Characteristics ofGeomaterials, 2003, pp. 293-298.
[5] Vuong TB, “Precision studies of resilient
modulus and permanent strain testing”, UnboundGranular Materials, 1999, pp. 63-83.
[6] Sawangsuriya A, “Assessment of Modulus and
CBR of Pavement Materials for Design andConstruction”, Bureau of Materials, Analysis
and Inspection, Department of Highways.
[7] Louay MN, Huang B, Puppala AJ, and Allen A,“Regression Model for Resilient Mokulus of
Subgrade Soils”, J. of the Transportation
Research Board, No. 1687, 1999, pp. 47-54.
[8] Semmelink CJ, and de Beer N, “Rapid
determination of elastic and shear properties ofroad-building materials with the K-mould, In
Unbound Aggregates in Roads ,University of
Nottingham, 1995, pp. 70-77.
[9] Eswards JP, Thom NH,Fleming PR, andWilliams J, “Axxelerated laboratory basemechanistic testing of unbound layers within the
mewly developed NAT Spingbox”, J. of
Transportation Research Board,2005, pp. 32-40.
[10] Thom NH, Cooper A, Grafton P, Ealker C, Wen
H, and Sha R, “A New Test for Base for BaseMaterial Characterisation”, Porceedings of theInternational Symposium on Heaby Duty
Asphalt Pavements and Bridge Deck Pavements,
2012.
[11] Koseki J, Mikami T, and Sato T, “Deformation
Characteristics of Granular Materials in Cyclic
One-Dimensional Loading Tests”, J. of
Transportation Infrastructure Geotechnology,
Volume 1, Issue 1, Jan. 2014. pp. 54-67.
[12] Nega A, Nikraz H, and AI-Qadi I, “Simulation of
Shakedown Behavior for Flexible Pavement’s
Unbound Granular layer”, J of American Societyof Civil Engineers, Jul 7 2015, pp. 801-812.
[13] Djellali A, Ounis A, and Saghafi B, “Behavior of
Flexible Pavements on Expansive Soils”, J of
Transportation Engineering, Volume 1, 2012.

CONCLUSION
The following conclusions can be derived fromthe test results presented in this study:
1. Coefficient of lateral earth pressure at rest canbe measured from the special CBR mould
attached with strain gauges for the measurementof tangential strain.
2. Constrain modulus of KMUTT sand, lateritic
soil, and crushed rock are not constant but
increases with an increase in the vertical stresslevel.
3. Constrain modulus that is free from bedding
error can be determined by performing tests
with different specimen’s heights. By using ananalytical method reported in the literature, it isfound that bedding error can result in asignificant underestimation of constrain
modulus.
ACKNOWLEDGEMENTS
The authors are grateful to The Thailand Research
Fund (TRF) via The Royal Golden Jubilee (RGJ) andKing Mongkut’s University of Technology Thonburi(KMUTT) for financial support (Contract number
PHD/0187/2556)
REFERENCES[1] AASHTO T307-99, “Standard method of test for
determining the resilient modulus of soils andaggregate materials”, American Association ofState Highway and Transportation Officials,
Washington, DC, USA., 2000.
[2] Chummuneerat S, Jitsangiam P, and Nikraz H,“Mechanical characteristics of hydrated cement
treated crushed rock base”, Advances inTransportation Geotechnics II, 2012, pp. 296301.
[3] Ibrahim SF, “Bagdad Subgrade Resilient
Modulus and Liquefaction Evaluation for
Pavement Design using Load cyclic Triaxial
Strenght”, J. of Environment and Earth Science,
Vlo. 3, 2003, pp. 125-141.

256

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

RATE-DEPENDENT DEFORMATION CHARACTERISTICS OF
EPS BEAD-MIXED SAND
Agawit Thaothip1 and Warat Kongkitkul2
Department of Civil Engineering, Faculty of Engineering,
King Mongkut’s University of Technology Thonburi, Thailand
1, 2

ABSTRACTThis research studies the deformation characteristics of expended polystylene (EPS) bead-mixed sand. A setof standard Proctor compaction test was conducted to determine the proper amount of water for use in the mixing
between EPS bead and sand at the volumetric ratio of 1:1. A special series of air-drained triaxial compression tests
was performed on the sand-EPS bead mixture (SEM) specimens. Various loading histories, which are: i)monotonic loading (ML) at constant strain rates; ii) sustained (creep) loading (SL); iii) cyclic loading with small
strain-amplitude (CL); and iv) stepwise changes in the strain rate during otherwise ML, were employed in thisstudy. These loading histories were used to evaluate the elastic and viscous properties of the SEM. Then theseproperties were also compared with the ones obtained from sand alone under similar loading conditions. It is foundthat the elastic and viscous properties of the SEM are qualitatively similar to those of sand alone. They are different
in terms of quantity. A non-linear three-component (NTC) model was used to simulate the elasto-viscoplastic of
the SEM. The simulations are well successful by using model parameters determined especially for the SEM
under the conventional model framework.
Keywords: Creep, Elastic, EPS bead, EPS bead-mixed sand, Non-linear three-component, Strain rate, Triaxial
compression test.
and iv) stepwise change in the applied shearing strain
rate, were used to evaluate strength properties, andelastic and viscous properties of SEM. Test results of
SEM are presented and compared with those of sandalone. Lastly, an elasto-viscoplastic non-linear threecomponent (NTC) model was used for simulations oftest results of SEM observed in the present study.

INTRODUCTION
Settlement of road embankment constructed on
the soft clay layer due to its primary consolidation isa crucial problem. This is in particular for the bridge
approach where differential settlement can result, andthus leads to danger for the drivers. At present, there
are many mitigation techniques proposed which canbe categorised into two groups; i.e., i) improvementof the soft clay layer by mixing with cement or otherrelevant additives [1]-[4]; and ii) reducing thesurcharge on the soft clay layer by using thelightweight fill materials [5], [6]. This researchinterests on the strength and deformation of propertiesof the latter group. A sand-expanded polystyrene(EPS) bead mixture (SEM) [7] was prepared bymixing at the ratio of 1:1 by volume, and therefore,
the unit weight of SEM became about a half of thesand alone. When used as a refilling material toreplace the original sand fill, the unit weight of SEM
is sufficient low so as to not allow the primaryconsolidation settlement of the soft clay to continue,
if the degree of consolidation of more than a certain
level had been achieved.
In view of the above, a series of special airdrained triaxial compression tests [8] were performedin this study. Various loading histories including: i)continuous monotonic loading with constant strain
rates; ii) sustained (creep) loading; iii) cyclic loading;

TEST MATERIALS
There are two test materials used to perform
triaxial tests in the present study. They are sand and
EPS bead.
Sand
A cleaned riverbed sand was used. To control its
gradation, the portion that passes through sieve No.
40 but retains on sieve No. 50 was mixed with theother portion that passes through sieve No. 50 but
retains on sieve No. 100 at the ratio of 1:1 by mass.
The gradation curve thus obtained is shown in Fig. 1.
EPS beads

A type of EPS bead was used. Its diameter isaround 1-2 mm. The absolute density is 32.9 kg/m3.

257

GEOMATE- Bangkok, Nov. 14-16, 2016

Sieve no.

200

100

100

50 40

16

loading (CL); and iv) stepwise change in the applied
shearing strain rate (SS). Triaxial compression test
program is shown in Table 1.

10

Table 1 Triaxial compression test program

Percent passing

80

40

D30 = 0.2272mm

20D10 = 0.1722mm

00.01

Fig. 1
1.00

d

3

Dry density,  (g/cm )

0.98

Test
S
E
W
SP
3 ε v 
SND01 875
163
ML
25
0.15
SND02 875
163
ML
50
0.15
0.15
ML
75
163
SND03 875
163
ML
100 0.15
SND04 875
0.15
ML
25
SEM01 550 12 105
0.15
ML
50
SEM02 550 12 105
0.15
ML
75
SEM03 550 12 105
100 0.15
ML
SEM04 550 12 105
100 0.05
ML
SEM05 550 12 105
SEM06 550 12 105 SL,CL 100 0.05
SEM07 550 12 105 SL,CL 100 0.05
100 0.50
ML
SEM08 550 12 105
SEM09 550 12 105 SL,CL 100 0.50
SEM10 550 12 105 SL,CL 100 0.50
SS100 0.05
SEM11 550 12 105SS
100 0.05
SEM12 550 12 105
Note: S = sand (g); E = EPS bead (g); W = water (g);SP = shearing patterns;  3 = confining pressure(kPa); v = basic vertical strain rate (%/min); SND =tests on sand alone; SEM = tests on sand-EPS beadmixtureD60 = 0.3216mm

60

0.1

Particle size in millimeters1

Particle size distribution of sand.
SEM

Maximum dry density = 0.97 g/cm
Optimum moisture content = 18.65%
3

0.96
0.94
0.92
0.90
0.88
10

Fig. 2

12

14

16

18

20

Moisture content, w (%)

22

24

TEST RESULTS AND DISCUSSIONS

Compaction curve by standard Proctor
compaction test on SEM.

Monotonic loading test results
From ML tests, deviator stress (q)-axial strain (a)characteristics between sand alone (Fig. 3(a)) andSEM (Fig. 3(b)) can be compared. It can be observed
from Figs. 3(a) and 3(b) that the q-a relations exhibit
a strain-hardening behaviour until achieving therespective peaks, and then show a strain-softening
behaviour toward the residual states. The behaviours
described above are qualitatively similar between thesand alone and the SEM specimens. This may imply
that sand in the SEM matrix play a major role tocontrol the global behaviours of SEM.
Shear strength parameters at the peak and theresidual states of sand alone and SEM are listed inTable 2. The peak and residual friction angles of sandalone are noticeably greater than the corresponding
values of SEM. On the other hand, cohesion at thepeak state of sand is smaller than that of SEM.

TEST METHODSStandard Proctor Compaction Test
The purpose of this test is to determine themaximum dry density and the optimum moisture
content of SEM. Fig. 2 shows the compaction curve
which defines the optimum moisture content (OMC)
of 18.65% and the maximum dry density of 0.97
g/cm3.
Air-Drained Triaxial Compression Tests
The purpose of this test is to evaluate the strengthand deformation properties of SEM. The specimen iscylindrical in the shape. It is 150 mm high by 70 mm
in diameter. Triaxial specimens were prepared bytamping the SEM layered in the mould such that thedensity meets the target value (0.97 g/cm3). Fourloading patterns were employed. They are: i)continuous monotonic loading with constant strain
rate (ML); ii) sustained loading (SL); iii) cyclic
258

GEOMATE- Bangkok, Nov. 14-16, 2016

400

400

Air-drained triaxial compression

Sand alone350

3=100 kPa

300

3=75 kPa200

3=50 kPa

100

Deviator stress, q (kPa)

Deviator stress, q (kPa)500

3=25 kPa

300
250
200
150
100

0

Deviator stress, q (kPa)

400

5

10

15

Axial strain, a (%)

SEM

400

Air-drained triaxial compression

300

350

3=100 kPa
3=75 kPa200

3=50 kPa100

3=25 kPa

00b)Fig. 3

 = 0.05%/min and 0.5%/min3 = 100 kPa

Stage 1

5

10

15

20

Axial strain, a (%)

25

30

Cohesion

Sand alone

Peak
Residual
Peak
Residual

(kPa)
5.0
0.0
16.5
0.0

SEM

200
150
100

Fig. 4

Table 2 Shear strength parameters of sand alone and
SEM

State

250

b)

35

Deviator stress (q)-axial strain (a) relations
from continuous ML triaxial compression
tests on: (a) sand alone; and (b) SEM

Sample300500Stage 3Cyclic loading (CL)Creep (SL)Monotonic loading (ML)

0

a)

20

Deviator stress, q (kPa)a)

End of loading

Stage1 q at 100kPa (R = 2.0)
Stage3 q at 200kPa (R = 3.0)
SL for 2 hr
CL 10 cycles q = 50 kPa

50

0

500

SL1-3

Elapsed time, t (not to scale)SL2-4

End of loading

Stage2 q at 150kPa (R = 2.5)
Stage4 q at 250kPa (R = 3.5)
SL for 2 hr
CL 10 cycles q = 50 kPa
Stage 4
 = 0.05%/min and 0.5%/min3 = 100 kPa

Stage 2Cyclic loading (CL)Creep (SL)Monotonic loading (ML)Elapsed time, t (not to scale)Loading histories used to evaluate elastic
modulus and creep behaviours of SEM at:(a) R =2.0 & 3.0 (SL1-3); and (b) R =2.5 &
3.5 (SL2-4).

clearly observed that upon the restart of ML after thelast CL at the highest deviator stress level, the q-arelation tends to rejoin to the relation that is obtained
from the respective continuous ML test, andimportantly, the peak deviator stress is maintained.
Thus it can be postulated that although creep strain of
SEM is obvious, it does not degrade the SEM’s shear
strength.
Unloading braches of CL of the last five loops
were exaggerated as shown in Fig. 6. To determine
the values of quasi-elastic Young’s modulus ( E eq )[9], lines were best-fitted to these respective
unloading branches. Then these slopes were averaged
and accounted as the representative of E eq for thedeviator stress level at which it was determined. Next
the E eq value was converted to be represented interms of R, where R is the stress ratio equal to σ1/σ3.
Fig. 7 shows the relation between E eq (expressed interms of R) and R, plotted in the full-log scale. It canbe seen that E eq is not constant but increases with anincrease in the stress ratio. For the same stress ratio,
the E eq of sand alone is noticeably greater than that ofSEM. The dependency of elastic modulus with thestress level observed with sand alone and SEM isqualitatively similar and can be expressed with hypoelasticity expressed by Eq. 1.

Friction
angle
(๐ )
43.1
42.0
36.0
34.5

Elastic modulus from CL tests
Two patterns of loading histories, used to evaluate
the elastic modulus and the creep behaviours of SEM,
are shown in Figs. 4(a) and 4(b). In these tests, ML
was firstly applied until the specified deviator stressat 100, 150, 200 or 250 kPa was achieved, and thensustained loading (SL) which lasted for two hours
was applied. Next cyclic loadings (CL) with a double
stress-amplitude of 50 kPa were applied for 10 cycles.
Two values of strain rate which are 0.05 and 0.5%/min were employed during ML and CL in theircourses (Figs. 4(a) and 4(b)). Test results thus
obtained are shown in Figs. 5(a) and 5(b) for the strain
rate of 0.05 and 0.5 %/min, respectively. It can be
259

GEOMATE- Bangkok, Nov. 14-16, 2016

100

450

Air-drained triaxial compression
SEM
350 Test name: E05-07
Strain rate during ML = 0.05 %/min
300 SL = 2 hr
CL 10 cycles q = 50 kPa
250
SEM07 SL2-4 (0)
3 = 100 kPa
200

SEM06 SL1-3 (0)

100

50

0

0

a)
500

Deviator stress, q (kPa)

400
350

4

250

6

Axial strain, a (%)

SL = 2 hr
CL 10 cycles q = 50 kPa
3 = 100 kPa

300

150

Fig. 7
SEM08 ML(100)

SEM10 SL2-4 (100)b)Fig. 52

4

Axial strain, a (%)

62600 = 0.05 %/min

250

q = 250kPa

240
230
220
210
200
3.50

3.52

3.54

Axial strain, a4Comparison of E eq  R relations between
sand alone and SEM
m(1)

Fig. 8 compares creep axial strain of SEM
developed after two hours between the sustained
loadings with the initial strain rates of 0.05 %/min and
0.5 %/min. The test results reveal that creep axial
strain increases with time. However, the creep strain
rate decreases with time towards nearly zero at theend of creep. Creep strain developed by sustained
loading for two hours increases with the stress ratio,
and at the same time, creep of SEM obtained with theinitial strain rate of 0.5 %/min is significantly larger
than the one with the initial strain rate of 0.05 %/min.
This shows the influence of strain rate at the start ofcreep on the development of creep strain with time.
In order to compare the creep axial strains having
different initial strain rates, they were corrected to theinitial strain rate of 3.5 x 10-4 %/min. Then thecomparison of creep axial strains between the sandalone and the SEM under the same initial strain rate
condition is shown in Fig. 9. The two different
relations for SEM shown in Fig. 8 are now becoming
close to each other in Fig. 9. The trend of increasingof creep axial strain with the increasing stress ratio is
obvious for both sand alone and SEM. However,creep of SEM is noticeably greater than the one ofsand alone as shown in Fig. 9.

3 = 100 kPa

190
3.48

3

Creep loading

8

Air-drained triaxial compression
SEM

270

2

Stress ratio, R=1/3

SEM, respectively; m = 0.576 and 0.485 for sandalone and SEM, respectively

Deviator stress-axial strain relations of SEM
obtained from tests with the intermissions ofSL and CL; the strain rate during ML and
CL is: (a) 0.05 %/min; and (b) 0.5 %/min
280

1where E 0  R %  = 7.20 and 5.66 for sand alone and500

= 0.05%/min)
SEM (

E eq =E 0  R 

SEM09 SL1-3 (100)

100

.
.
= 0.5%/min)
SEM (
.

Sand alone ( = 0.03%/min)

1

8

200

0

Deviator stress, q (kPa)2

Air-drained triaxial compression
SEM
Test name: E08-10
Strain rate during ML = 0.5 %/min

450Fig. 6

10

SEM05 ML (0)150

SEM & KMUTT sand alone

Eeq in terms R (R/%)

Deviator stress, q (kPa)

400

3.56

3.58

Unloading branches of the last five loops ofa CL course with the respective best-fittedlines to determine the quasi-elastic modulus

260

GEOMATE- Bangkok, Nov. 14-16, 2016

1.4

jump in the stress ratio upon a stepwise increase or
decrease in the strain rate was defined as shown in
Fig. 11. Then the measured stress ratio jumps werenormalised with the stress ratio immediately before
the jump (R/R), and then plotted against the ratio ofstrain rate after and before stepwise change, as shownin Fig. 12. The slope of the line best-fitted to the data
points shown in Fig. 12 (Eq. 2) is called the ratesensitivity coefficient () [10], [11].

Air-drained triaxial compression
SEM

Creep axial strain, a,cr (%)

1.2
1.0
0.8

0= 0.5 %/mino0.6

0= 0.05 %/min
o

0.4

ΔR

0.2
1.5

Fig. 80.8
0.7

Creep axial strain, a,cr (%)

0.6
0.5

2.0

2.5

3.0

Stress ratio, R = 1/3

3.5

R

4.0

Air-drained triaxial compression
SEM & Air-dried KMUTT sand alone
Creep strain rate 2 hours
for the same initial strain rate-4of 3.5x10 %/min

SEM, 0o

SEM, 100oNon-linear Three-component Model

0.3

A non-linear three-component (NTC) (Fig. 13)
was used to simulate rate-dependent characteristics(e.g., creep, jump in the stress ratio) of SEM. The
three components are elastic, inviscid, and viscouscomponents. Total stress ratio (R) consists of theinviscid and the viscous components (Rf and Rv)ewhile strain rate ( ε ) the elastic ( ε ) and theirirreversible ( ε ) components.
Elastic modulus (Eeq) of the elastic component,
which is dependent of stress ratio level (Fig. 7), is
determined by a hypo-elasticity model (Eq. 1). The
elastic strain rate can then be calculated from Eq. 3
[14].

0.2
0.1
0.0

Fig. 9Air-dried
KMUTT sand

2.0

2.5

3.0

Stress ratio, R = 1/3(2)The  value of SEM determined from this study is
0.0619. On the other hand, from previous studies onother standard sands by performing stepwisechangings in the strain rate during triaxial tests [12],[13], the  values are equal to 0.0226, 0.0195, and
0.0195 for Toyoura sand, Albany sand, and Monterey
sand, respectively. Therefore, SEM is more sensitive
to changes in the strain rate than typical sands forabout three times, which may due to the inclusion of
EPS bead into the matrix.

Comparison of creep axial strains of SEM
for two hours between the initial strain rates
of 0.05 %/min and 0.5 %/min

0.4

 ε irafter ir ε before =βlog10 

3.5

Comparisons of creep axial strains after
having been corrected for the same initialstrain rate of 3.5 x 10-4 %/min between sandalone and SEM

Strain rate responses of SEM

Responses of SEM to changes in the strain rate
were also studied in this research. To quantify theviscous responses, stepwise changes in the applied
shearing strain rate were performed. In these tests, thebasic reference strain rate ( ε 0 ) of 0.05%/min was
firstly selected. Different strain rates that are slower
and faster than ε 0 for 10 and 100 times and also theε 0 were then specified into a set. During a test, theshearing strain rate was stepwise changed many times
from a value to the other value specified in the set.
Fig. 10 shows relationship between the stress ratio
and the irreversible axial strain obtained with theabove-mentioned changings in the strain rate. It canbe readily seen that upon stepwise changes in thestrain rate, the stress-strain relation exhibited stressjumps. These stress jumps are responses due to theviscosity of SEM. To quantify the viscosity of SEM,eε = R E eq

(3)Inviscid stress ratio (Rf)-irreversible axial strain (irε a ) relation of the inviscid component, which is
independent of any strain rate effect and called thereference curve, is determined from the functional
form shown in Eq. 4.



R =P1 + P2 +P3 ε afir 1-exp -P  εira  -P Pa   (4)




4

P5

6

ε

ir
2



where P1 – P5 are constants that are determined fromthe regression analysis such that Eq. 4 is best-fitted toir
the respective R- ε a relation that is extrapolated tozero-strain rate.
261

GEOMATE- Bangkok, Nov. 14-16, 2016

θ= β r βθ ini +θ end
2+

θ ini -θ end

θ(ε )=θ end ; ε  ε θ
ir

ir

2

  ε   ir ir
cos  π  ir   ; ε <ε θ (6a)  εθ  
c

ir

(6b)

ir

ir

ir

ir

ir

v

{1-θ(ε )}  R
ir

ir

v
G.TESRAir

ir

R iso (ε ,ε )=R (ε )  g v (ε )
ir

(7b)

 dR  [r1 (ε ir )]ε -τ (7c)
(τ)

ir

v

R

ir

G.TESRAir(ε ,ε ,h s )=
ir

v

f

ir

ε

ir

τ=ε1

ir

g v (ε )=α  [1-exp{1-(
ir

ε

ir

ir
ε r

v

ir

iso

+1) }]
m

2

100

.

0

.

0/10

.

100

0

100

.

0/10

.

0

.

.

100

.

0

.

100

0/10

0/10

.

0 = 0.05 %/min

0/10
5

10

3 = 100 kPa

15

20

Irreversible axial strain, a (%)
ir

25

30

Fig. 10 Relationship between stress ratio andirreversible axial strain of SEM obtained
from a test with multiple changes in thestrain rate
5.0

Air-drained triaxial compression
Test name: SEM11

4.5

R

4.0

3.5

3.0

S2

S2

S1

.

step increare in
the strain rate
: strain rate S1<S2

R

S1

0 = 0.05 %/min3 = 100 kPa

12

15

step decreare in
the strain rate
: strain rate S1<S2
18

Irreversible axial strain, a(%)
ir

Fig. 11 Definitions of stress jumps upon a stepincrease and a step decrease in theirreversible strain rate(7a)

(ε ,ε ,h s )

3

.

.

0

0.2

0.1

R/R

v

100

.

.

.

.

1

where θ ini and θ end are the initial and residual values
irof transition of the  value; c and ε θ are constants.
v
The viscous stress ratio (R ) can be determined from
Eq. 7 [12].

R (ε ,ε ,h s )=θ(ε )  R iso (ε ,ε )+Air-drained triaxial compression
Test name: SEM11

4

Stress ratio, R = 1/3

θ(ε )=

5(5)

It was also found with SEM that, similar to manyother geomaterials [12], the  value is not constant but
dependent on the irreversible strain, and can be
expressed with Eq. 6.
ir6

Stress ratio, R = 1/3

The rate-dependent responses of SEM can beobserved not only with creep by sustained loading(Fig. 5) and stress ratio jumps by changes in the strain
rate (Fig. 10) but also with R-a relations bycontinuous ML tests with different strain rates asshown in Fig. 14. At the same a, it is obvious that R
for the faster strain rate is greater than the value forthe slower strain rate. This difference in the R value
along continuous ML R-a relations with different
strain rates is referred as residual state of ratedependency [12]. In Fig. 14, imaginary stress ratio
jumps were imposed between the two R-a relations,and then coefficient of rate-sensitivity at the residualstate (r) is then determined as shown in Fig. 12. Thenthe viscosity type parameter () [12] can then be
determined from Eq. 5.

0.0

-0.1

(7d)

-0.2
-3

where , m, ε r , and r1 are constants.
ir

SEM= 0.0619
r= 0.0585



= 0.0585/0.0619 = 0.95
r

-2

-1

0

1

2


r

Ratio of strain rate after and before stepwise change

3Fig. 12 Determination of rate-sensitivity coefficientfrom relationship between the normalised
stress ratio jump and the ratio of strain rate
after and before stepwise change262

GEOMATE- Bangkok, Nov. 14-16, 2016

6

SEM

Stress ratio, R

5Strain rate = 0.5 %/min

4Strain rate = 0.05 %/min

32

Experiment
Simulation

1

Fig. 13 Non-linear three-component model [10][12]0Simulation Results

5

10

Axial strain, a (%)

15

20

Fig. 14 Simulations of continuous ML tests on SEM
with different but constant strain rates of
0.05 and 0.5 %/min

Fig. 14 compares the R-a relations between theexperimental and simulation results from continuous
ML on SEM with strain rates equal to 0.05 and 0.5%/min. It can be readily seen that the NTC model canwell-successfully simulate all the R-a characteristics
that are different by the strain rate effects.
Fig. 15(a) compares R-a relations of SEM
subjected to SL during otherwise ML at the strain rateof 0.5 %/min between the experimental andsimulation results. It can be seen that not only thedevelopment of axial strain (creep) under a constantstress ratio condition but also the behaviour in that theR-a relation re-join to the one that would be obtained
from continuous ML can successfully be simulated.
Fig. 15(b) compares the creep axial strains fordifferent stress ratio values. The behaviours in that thecreep axial strain increases with an increase in the
stress ratio and also with an increase in the initialstrain rate at the start of creep can be well-simulated.
Thus the NTC can be used to predict the timedependent (rate-dependent) creep deformation of
SEM.
Fig. 16 compares R-a relations of SEM subjected
to stepwise changes in the strain rate between theexperimental and simulation results. All the sudden
increases and decreases of stress ratio by stepwise
increases and decreases in the strain rate can be wellsimulated. The NTC model therefore has a great
potential to predict the rate-dependent deformation
characteristics of SEM evaluated by the experimentsin the present study.

6

Stress ratio, R

5
4

SEM

creep 2 hr
cycle 10 cycles q = 50 kPa
0 = 0.5 %/min

Strain rate = 0.5 %/min3 = 100 kPa

3
2
Experiment
Simulation

1
0

a)
1.4

Creep axial strain,a,cr (%)

1.2
1.0

2

4

Axial strain, a (%)

At 2 hours

Strain rate (%/min)Experiment
Simulation

0.5

6

8

3.5

4.0

0.05

0.8
0.6
0.4
0.2

b)

1.5

2.0

2.5

3.0

Stress ratio, R = 1/3

Fig. 15 Simulations of 2-hr SL tests on SEM duringotherwise ML: a) comparisons of R-arelation with the strain rate during ML of 0.5%/min; and b) comparisons of creep axial
strain for different stress ratios

263

GEOMATE- Bangkok, Nov. 14-16, 2016

6

Stress ratio, R

5Air-drained triaxial compression & Simulation
Test name: SEM11

.

.

100

4

.

3

100

2

0

.

1
0

.

.

.

100

0

.

0/10

0/10

0/10

5

.

100

0

.

.

.

0/10

.

.

100

.

0

[2] Kawasaki T, Niina A, Saitoh S, Suzuki Y, andHonjo Y, “Deep mixing method using cement
hardening agent”, Proceedings of 10th
International conference on soil mechanics andfoundation engineering, Stockholm, 1981,
pp.721-724.
[3] Miura N, Koga Y, and Nishida K, “Applicationof a deep mixing method with quicklime for
Ariake-clay ground”, J. of Japan Soc. Soil Mech.
and Found. Eng’g, Vol.34 (4), 1986, pp.5-11.
[4] Terashi M, Tanaka H, and Okumura T.
“Engineering properties of lime treated marine
soils and deep mixing method”, Proceedings of
6th Asian regional conference on soil mechanics
and foundation engineering, Vol.1, 1979,
pp.191-194.
[5] Horvath J, “The compressible-inclusion functionof EPS geofoam: analysis and design
methodologies”, Res Rpt No CE/GE-98, 1998.
[6] NCHRP, Guideline and Recommended Standard
for Geofoam Applications in Highway
Embankments,
NCHRP
Report529,
Transportation Research Board, 2004.
[7] A. Edincliler , A. T. Ozer. “Effects of EPS beadinclusions on stress–strain behaviour of sand”. J.
of Geosynthetics International, Vol.21(2), 2014,
pp.89-102.
[8] Das, B. M. (2013). Triaxial test. In Fundamentalsof Geotechnical Engineer. Canada: Cengage.
[9] Kongkitkul W, Tabsombut W, Jaturapitakkul C,
and Tatsuoka F, “Effects of temperature on therupture strength and elastic stiffness ofgeogrids”, J. of Geosynthetics International,
Vol.19, No.2, April 2012, pp.106-123.
[10] Tatsuoka F, Ishihara M, Di Benedetto H, andKuwano R, “Time-dependent shear deformation
characteristics of geomaterials and theirsimulation”, J. of Soils and Foundations, Vol.42,
No.2, 2002, pp.103-129.
[11] Di Benedetto H, Tatsuoka F, and Ishihara M.:“Time-dependent
shear
deformation
characteristics of sand and their constitutive
modelling”, J. of Soils and Foundations, Vol. 42,
No.2, 2002, pp.1-22.
[12] Tatsuoka F, Di Benedetto H, Enomoto T,
Kawabe S, and Kongkitkul W, “Various
viscosity types of geomaterials in shear and theirmathematical expression”, J. of Soils andFoundations, Vol.48, No.1, February 2008, pp.
41-60[13] Kiyota T and Tatsuoka F, “Viscous property ofloose sand in triaxial compression, extension andcyclic loading”, J. of Soils and Foundations, Vol.
46, No.5, 2006, pp.665-684.
[14] Hoque E and Tatsuoka F, “Anisotropy in elastic
deformation of granular materials”, J. of Soils
and Foundations, Vol.38, No.1, 1998, pp.163179.

.

100

0/10

0 = 0.05 %/min3 = 100 kPa

10

15

Experiment
Simulation
Reference curve

20

Axial strain, a (%)

25

30

Fig. 16 Simulations of stepwise changes in thestrain rate applied to SEM
CONCLUSIONS
The following conclusions can be derived fromthe air-drained triaxial compression test results andtheir simulations.
1. Peak and residual friction angles of SEM are
noticeably smaller than those of sand alone.
2. Elastic Young’s moduli of air-dried sand alone
and SEM are not constant but increase with anincrease in the axial stress. At the same axial
stress, the elastic Young’s modulus of SEM isquite smaller than that of air-dried sand alone.
3. SEM exhibited creep strain significantly upon
the applied sustained loading. The creep strain
increases with an increase in the shear stresslevel and also increases with an increase in thestrain rate at the start of creep. Creep strain of
SEM is significantly larger than the one of sandalone. Yet, by ML at the end of creep, the stressstrain relation of SEM rejoins to the one that is
obtained by continuous ML with the same strain
rate, and thus, the peak shear strength ismaintained. Therefore, creep of SEM is not adegradation phenomenon.
4. SEM is more sensitive to the strain rate than the
air-dried sand alone. This also results in agreater amount of creep developed underotherwise the same conditions
5. NTC model can successfully simulate the stressstrain-time behaviours of SEM, taken into
account the dependency of elastic Young’s
modulus with the axial stress and the evaluatedand quantified viscous properties. Moreover,this model can successfully predict the creep
strain developed during sustained loading.
REFERENCES[1] Chida S, “Dry jet mixing method-State of the Art
on Improvement Methods for Soft Ground”, J. of
Japan Soc. Soil mechanics and foundationengineering., 1982, pp.69-76,264Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

GEOTECHNICAL REQUIREMENTS FOR CAPTURING CO2
THROUGH HIGHWAYS LAND
M. Ehsan Jorat1, Ben W. Kolosz1, Mark A. Goddard1, Saran P. Sohi2, Nurten Akgun1, Dilum Dissanayake1
and David A.C. Manning1
1
School of Civil Engineering and Geosciences, Newcastle University, UK;
2
School of Geosciences, University of Edinburgh, UK

ABSTRACTRoadside verges in Britain support 238,000 hectares of vegetated land and approximately 10 hectares ofvegetated central reserves. These areas have the potential to be engineered in such a way that they deliver arange of ecosystem services such as flood regulation and biodiversity conservation in addition to their primaryfunctions such as comfort of sidewalk users (mostly un-vegetated), protection of spray from passing vehicles, aspace for benches, bus shelters, street lights and other public amenities, and visual improvement of the roads anddesignated green belts. Previous research has shown that in soils, calcium-rich materials such as recycled
crushed concrete or crushed dolerite undergo carbonation. This effectively captures CO 2 from the atmosphereand stores it in the form of CaCO 3 precipitated between soil particles. Engineering this process can potentiallyassist the UK in achieving its ambitious target to reduce CO 2 emissions by 80% of 1990 levels by 2050. Rates ofcarbonation measured at urban brownfield sites in the UK suggest that treating 12,000 ha of land containing
suitable amendments could remove 1M t CO 2 annually. However, brownfield sites are often subjected to reconstruction activities which would reduce the rate of CO 2 absorption from the atmosphere by sealing. To
optimize the rate of carbonation, engineered soils need to be constructed at locations subjected to least postconstruction activities and roadside verges and central reserves represent a key opportunity in this regard. This
paper calculates limits to CaCO 3 formation within the first 1 m of pore spaces of soils at roadside verges andcentral reserves in Britain considering soil porosity of 20%.
Keywords: Britain roads, engineering soil, CO 2 capture, roadside verges and central reserves
However, capturing and storing atmospheric carbon
in a form of carbonate in urban soil also appears to
be an economically and environmentally effective
method [5], [6], [7]. Measurements at an urban
brownfield site in the UK suggest that in a few years,
each 12,000 ha of engineered land could remove 1
Mt CO 2 yr-1 [7]. The formation of pedogenic
carbonates, which are predominately composed ofthe mineral calcite (CaCO 3 ), depends on two main
factors (i) the availability of calcium; and (ii)carbonate in solution.
Calcium is naturally derived from the weatheringof silicate minerals (plagioclase feldspars, pyroxenesetc.) that commonly occur in basic igneous rocks(e.g. basalts and dolerites), or from artificial calcium
silicate and hydroxide minerals within concrete andcement [8], [9]. Carbonate carbon is predominantly
derived from photosynthesis, based on stable isotope
studies [10]. During the process, silicate minerals, inparticular calcium silicates, react with dissolved
CO 2 to form carbonates (Equation 1).

INTRODUCTION
Roadside verges and central reserves are an integral
part of road construction. Pieces of land between theroad-surface and the boundary line, e.g. adjacenthedgerows, fences or hard development, are called
roadside verges [1]. Central reserves are the areasthat separates the carriageways of a dual
carriageway exclusive of any hardstrips [2].
There are estimated to be 238,000 hectares ofroadside verges in Britain [3]. In 2014, the United
Kingdom’s Department of Transport estimated thelength of roads (Motorways and type A, B, C and U
roads) in Britain to be 245,800 miles. Considering
the length of roads and central reserves associatedwith it, there are nearly 31,200 miles of central
reserves along the roads in Britain.
Roadside verges and central reserves are areas ofgreat potential to capture carbon in addition to theirprimary functions, if managed and designed
appropriately. The ambitious UK target of reducing
CO 2 emissions by 80% by 2050, compared to a1990 baseline, can only be achieved by engaging
diverse sectors of society. It is likely that major CO 2
emission reduction will be achieved throughimproved efficiency in the energy system (e.g. [4]).

CaSiO 3 + CO 2 + 2H 2 O
(1)

265

CaCO 3 + H 4 SiO 4

GEOMATE- Bangkok, Nov. 14-16, 2016

This leads to the removal of atmospheric carbon
and its subsequent storage in a stable, inorganic state[6], [7], [8], [11], [12], [13], [14], [15]. Once
formed, this is a stable store for CO 2 since it will
only be naturally removed by dissolution, whencarbon would normally stay stable in solution,entering surface and groundwater systems.
In this paper, we investigate possible constraints
to realising the potential for CaCO 3 sequestration inroadside verges and central reserves of Britain. The
result of this investigation will assist in designing
more sustainable roads which would provide theirprimary functions while also sequestering CO 2
emissions discharged into the atmosphere partly
from vehicles using internal combustion engines.
We also highlight the implications of engineering
roadside verges and central reserves to capture
carbon on their ability to also deliver a range ofadditional ecosystem services, such as flood
regulation and biodiversity.
THEORETICAL BACKGROUNDCaCO 3 formation occurs within soil pore spaces.
The porosity of soils ranges between 20–60%,
depending on soil type and degree of compaction[16]. Pore spaces within coarser soils provide higher
potential for carbon sequestration because larger
pore spaces having lower surface areas result in
decreased friction and provide higher permeability.
Higher permeability results in faster weathering ofsilicate minerals and provides sufficient H 2 O for thechemical reaction (Equation 1) to take place.
For roadside verges and central reserves toprovide their primary functions, they should reach acertain degree of compaction that leads to significant
reduction of porosity. However, optimum
compaction of soil with small grains fitting within
the larger grain pores would give a porosity ofaround 20% [16]. This means that even at thehighest degree of compaction, 20% of the total soilvolume is comprised of voids which can provide a
potential location for CaCO 3 formation.
Based on previous measurements by [6] at the
Science Central development site in Newcastle, UK,
the majority of carbonation occurs in the top 1 m ofthe soil profile, depending on soil type andprecipitation. Based on that result, the first 1 m soil
depth is also considered for roadside verges andcentral reserves. The porosity of the soil at roadsideverges and central reserves is considered to be 20%.

FIG. 1 Dimensions of cross-section components for(a) rural motorway, (b) urban motorway, (c) rural
single carriage all-purpose, (d) rural dual carriage
all-purpose, (e) urban single carriage all-purpose and(f) urban dual carriage all-purpose roads mainline[2].
Based on an estimated density of pure CaCO 3 as2715 kg/m3 [17] for soil at roadside verges andcentral reserves, maximum ranges of CaCO 3
deposition within pore spaces are calculated andpresented in Table 2. For the calculations, theporosity of 20% is considered and it is assumed that
100% of the pore spaces are filled with CaCO 3 .
Therefore, potentially 543 kg of CaCO 3 for each
cubic metre of soil could be stored within soil porespaces in roadside verges and central reserves.

RESULTS
Cross-sections and dimensions for rural and urban
roads in Britain are shown in Fig. 1. Based on thedimensions provided, the total volume of the soil
associated with the first 1 m of road central reserves
is calculated and presented in Table 1.
266

GEOMATE- Bangkok, Nov. 14-16, 2016

Clearly, engineering all roadside verges andcentral reserves for carbon sequestration in Britain is
not feasible since they are mostly operational andreconstruction to implement carbon capture design
into the roadside verges and central reserves requires
significant funding and time. However, Departmentof Transport statistics in 2014 show that total road
length in Britain increased 2% over ten years,
compared with 2004 baseline [18]. Considering asimilar rate for the next ten years, if roadside verges
and central reserves of new roads are engineered forcarbon capture purposes during construction they
have the potential to deposit nearly 25 MT of
CaCO 3 which equates to 11 MT of CO 2 . This figure
is separate from roads which will be upgraded andresurfaced in future providing further opportunity to
implement the carbon capture design during reconstruction. The calcium source required for newengineered soils could be partly acquired throughrecycling of concrete from the demolition of bridges
and other concrete structures associated with roads
as they reach the end of their life. Remaining
calcium sources could be acquired from local
demolition sites or dolerite quarries.
Green infrastructure needs to be multifunctional(e.g. [19]) and carbon capture is only one of anumber of ecosystem services that roadside verges
and central reserves can provide. Although roads
generally have a negative impact on animal
populations (e.g. [20], [21]), sympatheticallymanaged or restored roadside verges make animportant contribution to the conservation ofgrassland species (e.g. [22]) and pollinating insects[23]. For instance, Highways England are committed
to the implementation of the National Pollinator
Strategy for England [24] and will undertake aprogramme of works to achieve a significant area ofspecies rich grasslands estimated at 3500 hectares by
2021.
The choice of planting and management ofvegetation on roadside verges and central reserves
could also influence calcite formation. Carbon storedin carbonates is plant derived, implying thatencouraging plant growth on engineered urban soils
may facilitate carbon capture [8], as well as
providing other ecosystem services associated withvegetated land.
If fully carbonated, the permeability of soil
approaches zero and roadside verges would nolonger have the capacity to drain water, possibly(depending on specific design) leading to flooding
during periods of heavy rainfall. For roads in generaland motorways in particular, if it occurs, suchflooding might lead to damaging events.
Accordingly, a proper drainage system such as
SUDS [25], [26] should be considered for roadsideverges and central reserves if engineered to capture
carbon in addition to their primary functions. A

Table 1 Total volume of soil present within the first
1 m of soils along road central reserves in Britain
Type of roads

Rural/urban
motorway
Rural dual
carriage
Urban dual
carriage
Rural/urban
dual
carriageway

Total
length ofcentralreserve(mile) [18]Volume of soilpresent in theupper 1 m ofcentral reserves
(m3)

2212

8815

22122

88710

6882

19870

Not
available inurban roads

Not available inDual carriage
roads

Table 2 Total potential CaCO 3 storage capacity
within the top 1 m of soil depth along roadsideverges and central reserves in Britain. For the
calculations of this table, porosity of 20% is
considered
Type

RoadsidevergesRoad central
reserves

Volume of soilpresent in the
top 1 m (m3)
2,380,000,000 1

Potential
CaCO 3storage (Mt)

117,395

0.06

1292

∑ = 1292

DISCUSSION
Results from Table 2 show that roadside verges
have a great potential for CaCO 3 deposition
compared with road central reserves which devote avery small and negligible portion to potential CaCO 3
deposition.
Table 2 indicates the limit to the formation of
CaCO 3 within the first 1 m of sediments at roadsideverges and central reserves, which is defined byporosity and hence particle size and degree ofcompaction. Considering the porosity of 20% andassuming all of the void spaces within the top 1 m ofsoil within roadside verges and central reserves arefilled with CaCO 3 , they have the capacity to deposit
an estimated maximum of 1,292 Mt CaCO 3 (Table2), equivalent to 150 Mt C. However, in reality,
other factors such as solution pH, the presence oforganic complexes and fluid flow rate also play animportant role in the formation of CaCO 3 [11].

267GEOMATE- Bangkok, Nov. 14-16, 2016

further benefit of SUDS is that the created wetland
can provide habitat for supporting biodiversity.
Predicting the saturation point for the CaCO 3
formation between soil particles is an essential factor
which largely controls the geotechnical properties ofthe substrate such as permeability and this can bedone using multiple regression modelling. Throughregression analysis, quantitative dependent variable
distribution would be created based on one or morequantitative independent variables [27] which take
into account the conditional distribution of thedependent variable as a function of independent
variables.
The continuous quantitative dataset of possible
independent predictors which could be used in themultiple regression model include substrate porosity,
grain size, degree of compaction, permeability, Ca
content, organic content, pH and vehicle traffic flow.
Using the independent variables, the dependent
variable can be defined. This will be used for theprediction of CaCO 3 formation in the soil. The
model would provide an understanding of thesaturation point in which the engineered soil atroadside verges stops sequestering CO 2 .
By adding significant variables in a regression
model, the accuracy of the dependent variables will
increase. However, overloading the model withexcessive variables could lead to inefficiency of themodel and result in overfitting of the results. In
contrast, having only a few variables would result inbiased results. To determine the risk profiles of thevariables, which are used in the multivariable model,
the effective index is commonly used in the
regression models [28].
The dependent and independent variables’ lower
limits need to be specified in order to develop an
accurate prediction by a multiple regression model.
The minimum sample size is determined by the
relationship between the number of independent
variables and the squared multiple correlation
coefficient (R2) [29]. Determining the sample size
increases the accuracy of results and will facilitate
progress planning and cost management.

associated with new roads have the theoretical
capacity to store 25 Mt CaCO 3 (from 11 Mt CO 2 ; 3
Mt C). Recycling of concrete from demolition ofbridges and other concrete structures associated withroads as they reach the end of their life could be an
appropriate calcium source for new engineered soil.
This study indicates that road components such as
roadside verges and central reserves have potential
to provide the function of passive CO 2 removal.
However, considering the possibility that depositionof CaCO 3 within soil pore spaces may reduce
permeability, efforts to maintain proper drainage inthe soil profile may be required to avoid flooding.
Engineering a carbon capture function into roadsideverges needs to be sympathetic to the provision ofother ecosystem services, such as biodiversity
conservation.
ACKNOWLEDGMENTS
The authors thank funding of the Engineering andPhysical Sciences Research Council (EP/K034952/1
and EP/I002154/1). We thank the United Kingdom
Department of Transport for providing the necessary
resources and information.
REFERENCES
Thomas C, “ROADSIDE VERGES”,
Warwickshire, Coventry and Solihull Local
Biodiversity Action Plan, 2005, pp.8.
[2] Department for Transport, “Design Manual for
Roads and Bridges - Cross-Sections andHeadrooms”, Volume 6, Section 1, Part 2, 2005.
[3] Department for Transport, “Road lengths in
Great Britain”, 2013, available from:https://www.gov.uk/government/uploads/system/uploads/attachment_data/file/208692/roadlengths-in-great-britain-2012.pdf.
[4] Pacala S, Socolow R, “Stabilization wedges:
Solving the climate problem for the next 50
years with current technologies”, Science 305,
2004, pp. 968– 972.
[5] Manning DAC, “Biological enhancement of soil
carbonate precipitation: passive removal of
atmospheric CO 2 ”, Mineralogical Magazine 72,
2008, pp. 639-649.
[6] Washbourne CL, Renforth P, Manning DAC,
“Investigating carbonate formation in urban
soils as a method for capture and storage ofatmospheric carbon”, Science of the Total
Environment 431, 2012, pp. 166-175.
[7] Washbourne CL, Lopez-Capel E, Renforth P,
Ascough PL, Manning DAC, “Rapid removal ofbyurban
soils”,
atmospheric
CO 2
Environmental Science and Technology 49,
2015, pp. 5434-5440.
[8] Manning DAC, Renforth P, “Passive
sequestration of atmospheric CO 2 through[1]

CONCLUSIONS
Limits to CaCO 3 formation within pore spaces ofsediments at roadside verges and central reserves
have been calculated and we show that roadsideverges have the greatest potential deposition rate.
Considering a porosity of 20%, the top 1 m of soil atroadside verges and central reserves along rural andurban roads in Britain has a theoretical capacity of
1,292 Mt CaCO 3 , equivalent to 150 Mt C, of whichalmost all is in verges. Road length in Britain is
expected to increase by 2% over the next ten years.
This will lead to increase vehicle movements andCO 2 emissions. If managed and engineered
appropriately, roadside verges and central reserves
268

GEOMATE- Bangkok, Nov. 14-16, 2016

coupled plant-mineral reactions in urban soils”,
Environmental Science and Technology 47(1),2013, pp. 135-141.
[9] Jorat ME, Goddard MA, Kolosz BW, Sohi S,
Manning DAC, “Sustainable Urban Carbon
Capture: Engineering Soils for Climate Change(SUCCESS)”, In: 16th European Conference on
Soil Mechanics and Geotechnical Engineering(XVI ECSMGE 2015). 2015, Edinburgh, pp.
2559 – 2564.
[10] Cerling TE, “The stable isotope composition ofmodern soil carbonate and its relationship toclimate”, Earth and Planetary Science Letter
71,1984, pp. 229-240.
[11] Renforth P, Manning DAC, Lopez-Capel E,“Carbonate precipitation in artificial soils as asink for atmospheric carbon dioxide”, Applied
Geochemistry 24, 2009, pp. 1757-1764.
[12] Renforth P, Edmondson J.; Leake, J. R.; Gaston,
K. J.; Manning, D. A. C. Designing a carbon
capture function into urban soils. Proc. Inst. Civ.
Eng.: Urban Design Plann. 2011, 164 (2), pp.
121−128.
[13] Kolosz BW, Goddard MA, Jorat ME, Aumonier
J, Sohi SP, Manning DAC, “A Sustainability
Framework for Engineering Carbon Capture
Soil In Transport Infrastructure”, International
Journal of Transport Development andintegration 2017, 1, pp. 74-83 (In-press).
[14] Kolosz BW, Manning P, Goddard MA, Jorat
ME,
Sohi
SP,
Manning
DAC,“A
Multifunctional
Soil
Inorganic
CarbonPrediction Model”, In: 8th International
Congress on Environmental Modelling and
Software. 2016, Toulouse France (In-press).
[15] Kolosz BW, Goddard MA, Jorat ME, Sohi S,
Manning
DAC,“Developing
Lifecycle
Inventory Indices for Estimating the Carbon
Sequestration of Artificially Engineered Soils
and Plants”, In: 5th Asian conference on
Sustainability, Energy, and the Environment,
2015, pp. 93-112.
[16] Jorat ME, Kolosz BW, Sohi S, Lopez-Capel E.,
Manning DAC, “Changes in geotechnicalproperties
of
urban
soils
duringcarbonation”, In: 15th Pan-American Conference
on Soil Mechanics and GeotechnicalEngineering, 2015, Buenos Ares, Argentina, pp.
912 – 918.
[17] Deer WA, Howie RA, Zussman J, “Anintroduction to the rock forming minerals”,
Longman Scientific & Technical, Essex, 2013,
3rd Edition.
[18] Department for Transport, “Road lengths in
Great Britain”, 2014, available from:https://www.gov.uk/government/uploads/system/uploads/attachment_data/file/428857/roadlengths-in-great-britain-2014.pdf.

[19] Lovell ST, Taylor JR, “Supplying urban
ecosystem services through multifunctional
green infrastructure in the United States”,
Landscape Ecology, 28, 2013, pp. 1447-1463.
[20] Muñoz P, Torres F, Megías A, “Effects of roads
on insects: a review”, Biodiversity andConservation, 24, 2015, pp. 659-682.
[21] Benítez-López A, Alkemade R, Verweij PA,“The impacts of roads and other infrastructure
on mammal and bird populations: A metaanalysis”, Biological Conservation, 143, 2010,
pp. 1307-1316.
[22] Auestad I, Rydgren K, Austad I, “Road verges:
Potential refuges for declining grassland species
despite remnant vegetation dynamics”, Annales
Botanici Fennici, 48, 2011, pp. 289-303.
[23] Hopwood JL, “The contribution of roadsidegrassland
restorations
tonativebee
conservation”, Biological Conservation, 141,2008, pp. 2632-2640.
[24] DEFRA, “The National Pollinator Strategy: forbees and other pollinators in England”, 2014.
[25] Poleto C, Tassi R, “Sustainable Urban Drainage
Systems”, Drainage Systems, Prof. Muhammad
Salik Javaid (Ed.), ISBN: 978-953-51-0243-4,
InTech, 2012, available
from: http://www.intechopen.com/books/draina
ge-systems/sustainable-drainage-systems.
[26] Zhou Q, “A Review of Sustainable Urban
Drainage Systems Considering the Climate
Change and Urbanization Impacts”, Water 6,
2014, pp. 976-992.
[27] Fox J, “Applied Regression Analysis, Linear
Models, and Related Methods”, Published bySage Publications, Inc., ISBN: 0-8039-4540-X,
the United Stated of America, 1997, Chapter 2,
pp.15.
[28] Harrell F E Jr, “Regression Modelling
Strategies: With Application to Linear Models,
Logistic Regression and Survival Analysis”,
published by Springer Science Business Media
ISBN-10: 0-387-95232-2, United States ofAmerica, 2011.
[29] Knofczynski GT, Mundfrom D, “Sample Sizes
When Using Multiple Linear Regression for
Prediction”, published by Educational andPsychological Measurement, 68(3), 2008, pp
431-442.

269

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

COMPOSITIONAL CHARACTERISTICS OF SAND MATRIX SOILS
Choy Soon Tan1, Aminaton Marto2 and Ali Sobhanmanesh1
Faculty of Engineering, Technology and Built Environment, UCSI University, 56000 Cheras, Malaysia
2
Faculty of Civil Engineering, Universiti Teknologi Malaysia, 81310 UTM Skudai, Malaysia

1

ABSTRACTPrevious study always assumes that the typical engineering characteristic of clean sand was remarkably
representing as to all types of sand matrix soils. However, the primary physical characteristics are varying withthe amount of presenting fine particles within the matrix of sand grain. Empirical evidences revealed that thesand matrix soils are also liquefiable besides the clean sand. Hence, the typical engineering characteristic ofclean sand could not represent as to all types of sand matrix soils. The aim of this paper is to examine theprimary physical characteristics of sand matrix soils through experimental study.
Keywords: particle size distribution, particle density, limiting density, void ratio
INTRODUCTION

sand matrix soils gives significant influences to thecompositional characteristic. It is believed that thisis the main reason why current empirical evidences
show that sand with limited percentages of fines isalso liquefiable. Numerous researchers initiated theirresearch to characterize on which type of sandmatrix soils is the most susceptible to soilliquefaction. The boundary limit on liquefaction
susceptibility criteria have been broadened sincethen. Therefore, it is essential to be able to describe
the physical characteristics of soils more concisely
in order to express the real engineering behaviour ofsand matrix soils [6]. The plasticity behaviour ofsand matrix soils have been described in companion
paper of [7]. Therefore, the aim of this paper is to
examine the primary physical characteristics of sandmatrix soils through experimental study. The
primary physical characteristics are the particle
density (ρ s ), particle size distribution and limiting
densities (ρ max /ρ min ).

Chinese Criteria [1] is the best known
liquefaction susceptibility identification founded
following the 1975 Haichang Earthquake and the1976 Tangshan Earthquake. The criteria are: (1)fraction finer than 0.005mm ≤ 15%; (2) liquid limit≤ 35% and (3) natural water content ≥ 0.9 liquid
limit. Some modification and amendment have been
done over the years. Seed et al. [2] combined thecriteria into a plasticity chart while Finn [3]suggested some adjustments to suit globally
recognised ASTM standard. The Modified Chinesecriteria are now the basic groundwork ofliquefaction susceptibility criteria.
In the 20th century, researchers started toquestion and debate on the usability of Modified
Chinese Criteria owing to the presence of fines(which supposed to be non-liquefiable) in some field
evidences which deviated from the criterion. Bothacademicians and engineers believed that theModified Chinese Criteria that is conceptually basedon empirically oriented, which is inadequate todefine the vulnerability of sand with some fines
towards soil liquefaction. Thus, the re-examination
is vital to introduce some necessary correction
factors in the criteria in order to adjust in accordancewith the influence of fines [4].
Kramer [5] reviewed and concluded that there
are four important criteria in liquefaction
susceptibility which are historical criteria, geologic
criteria, compositional criteria and state criteria.
Compositional characteristic is the most importantcriteria out of the other three criteria because it isassociated with volume change behaviour. These
compositional characteristics include particle size,
shape and gradation. The presence of plastic fines in

EXPERIMENTAL TESTING
In order to quantify the physical characteristics
of various kinds of sand matrix soils, thereconstituted sand matrix soils were mixed byadding two types of commercially available plastic
fines to the parent sand at various percentages byweight. The clean sand used to form the sand matrix
soils in this study is the natural sand obtained from ariver in Johor Bahru, Malaysia. The plastic fines are(i) low plastic fines, white kaolin (MI) by Kaolin(Malaysia) Sdn Bhd and (ii) highly plastic fines,
green bentonite (CE) by Halliburton (Malaysia) Sdn
Bhd. Four main types of reconstituted sand matrix
soils was summarised in Table 1.

270

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Compositional percentages of reconstituted sand matrix soils

Sand-Bentonite
Mixtures

Sand-Fines
Mixtures

Sand-Kaolin
Mixtures

Clean
Sand

Sand
Matrix
Soils

Percentages by weight (%)
Label

Sand

KaolinBentonite

100

0

0

SAND100

95
90
85
80
75
70
60
90
80
80
80
75
70
60
95
90
85
80
75
70
60

5
10
15
20
25
30
40
5
15
10
5
12.5
15
20
0
0
0
0
0
0
0

0
0
0
0
0
0
0
5
5
10
15
12.5
15
20
5
10
15
20
25
30
40

SK05B00
SK10B00
SK15B00
SK20B00
SK25B00
SK30B00
SK40B00
SK05B05
SK15B05
SK10B10
SK05B15SK13B12SK15B15
SK20B20
SK00B05
SK00B10
SK00B15
SK00B20
SK00B25
SK00B30
SK00B40

The soil classification test that was carried out inthis study to investigate the physical characteristics
of sand matrix soils are (i) particle density test using
the small pycnometer method; (ii) particle size
distribution test by combining the results of wet
sieving method and hydrometer test and (iii) limiting
density test. All tests were carried in accordancewith British Standard (BS) 1377-2:1990 [8] except
for the limiting density test. In fact, the limiting
densities of sand should be determined inaccordance with Clause 4.2 for a maximum densityof sands and Clause 4.4 for a minimum density ofsands in BS1377-4:1990 [9]. However, the general
requirement in both clause states that both methods
is suitable only for sand containing a small amountof fine material, usually not exceeding 15 % ofmaterial finer than a 63µm test sieve, by mass. Dueto that the soil sample used in this study, the sandmatrix soils are mostly having more than 10 % offines by weight based on the results of particle size
distribution; the techniques recommended by [10]were used to determine the limiting densities.

pycnometer method is more suitable for the soils
fines than 2 mm, as chosen to be conducted in thisresearch. All soil samples with approximately 100 gwere oven dried at 105 °C prior to testing. The soil
was divided into three portions to carry out the test
concurrently to ensure the repeatability andprecision of the results. The test was carried out inaccordance with Clause 8.3 in Part 2 BS1377.
Particle Size Distribution
Soil consists of individual particles or grains.
Grain size refers to the size of an opening in a squaremesh through which a grain will pass. Since all of
the grains in a mass of soil are not the same size, it isconvenient to quantify grain size in terms of a
gradation curve. The particle size distribution is thepercentages of the various grain sizes present in asoil. Since the soil samples in this study are ranging
from sand size down to clay soils, the particle size
distribution need to be determined by combining theresults of wet sieving method and the results of thesedimentation test by using the hydrometer method.
All soil samples were oven dried at 105 °C prior to
testing. The wet sieving method was carried out inaccordance with Clause 9.3 in Part 2 BS1377 while
the hydrometer method was carried out inaccordance with Clause 9.5.

Particle Density
Particle density of the soil, G S is the ratio of themass density of the mineral soils in soil normalised
relatively to the mass density of water. The small

271

GEOMATE- Bangkok, Nov. 14-16, 2016

Limiting Density
As the clean sand used in this study is the riversand, and hence, the presence of other organic
material in the sand grain causes a lower value ofparticle density in this study. Both the plastic fines
also reported a lower particle density than the typicalvalues. The reason is because the manufacture
process of these commercially available fines had
eliminated the presence of heavy metals, in order tokeep the purity of the material. In fact, particle
density is remarkably influenced by its mineralogy
composition while the natural soil often consists of avariety of minerals. Therefore, chemical tests which
are beyond the scope of this study should be
conducted in the future to determine the mineralogyof the clean sand and both plastic fines.

The limiting density is the dry densities
corresponding to the extreme states of packing ofparticles of a granular soil. The maximum density isdefined as when the soils at the densest practicablestate of packing while the minimum density is when
the soils at the loosest state of packing. In fact,
neither British Standard (BS) nor American Societyof Testing Material procedure (ASTM) is suitable
for the determination of the maximum density andminimum density for sand matrix soils; moststandard are limited to soils with maximum fines
content of 10% to 15% only. Due to that the soil
sample used in this study, the sand matrix soils aremostly having more than 10% of fines by weight;
the techniques recommended by Lade andYamamuro [10] were used to determine the limiting
densities in this study. The soil samples were ovendried at 105 °C prior to testing.

Particle Size Distribution
The sieving method is not practical for soils
smaller than silt size. By combining the results ofwet sieving and sedimentation procedure usinghydrometer, the particle size distribution of thewhole sand matrix soils is achievable. The grading
curve from sand size down to clay size is shown in
Fig. 2. The composition percentages of the sandmatrix soils are listed in Table 2 together with its
grading characteristics, including the coefficient of
uniformity, C u and coefficient of curvature, C c . The
sand matrix soils were then classified in accordancewith the British Soil Classification System forEngineering Purposes (BSCS) as described in clause
41 to Clause 43 of BS 5930:1981 [11]. Althoughmost of the classification tests in this study arecarried out in accordance with BS, the classification
based on the Unified Soil Classification System(USCS) was adopted in this study for comparison
purpose. The USCS [12] is more commonly used in
the industry to describe the texture and grain size ofmost unconsolidated materials. The composition
characteristics of sand matrix soils at various
percentages of fines content are more obvious byusing the USCS.

RESULTS AND DISCUSSION
The results of the soil classification test to
describe the primary physical characteristics of sandmatrix soils are presented in this section. The
secondary physical characteristics which are
obtained by correlate with the primary physicalcharacteristics are also included.
Particle Density

Particle Density,ρs(Mg/m3)

Particle density (G S ) is typically expressed using
three significant figures. The G S for sands is often
assumed to be 2.65 and higher for the clay, normally
between 2.70 to 2.80 depending on mineralogy.
However, the pure soil used in this research has
lower value of G S compared to the typical values.
The G S value for the clean sand is 2.63 while the G S
for the kaolin and bentonite are 2.61 and 2.35
respectively. The results of G S values for sandmatrix soils are shown in Fig. 1.
2.65
2.60
2.55
2.50
2.45
2.40
2.35
2.30
0
Sand-Bentonite

5 10 15 20 25 30 40 100
Fines Content, FC (%)
Sand-Kaolin-Bentonite

Sand-Kaolin

Fig. 1 Particle density of sand matrix soils

272

GEOMATE- Bangkok, Nov. 14-16, 2016

Percentages Passing (%)

100
80

SK05B00

60

SK10B00
SK15B00

40

SK20B00

20

SK25B00
SK30B00

0
0.001

0.010

0.100
1.000
Particle Size (mm)10.000SK40B00(a) Sand-kaolin mixtures

Percentages Passing (%)

100
80

SK05B05

60

SK15B05
SK10B1040

SK05B15

20

SK13B12SK15B15

0
0.001

0.010

0.100
1.000
Particle Size (mm)10.000SK20B20(b) Sand-fines mixtures

Percentages Passing (%)

100
80

SK00B0560

SK00B10
SK00B1540

SK00B20

20

SK00B25
SK00B30

0
0.001

0.010

0.100
1.000
Particle Size (mm)10.000

SK00B40(c) Sand-bentonite mixtures
Fig. 2 Particle size distribution of (a) sand-kaolin mixtures pure soils; (b) sand-fines mixtures; (c) sand-bentonite
mixtures273

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 Grading Characteristics of sand matrix soils

SeriesSAND100
SK05B00
SK10B00
SK15B00
SK20B00
SK25B00
SK30B00
SK40B00
SK05B05
SK15B05
SK10B10
SK05B15SK13B12SK15B15
SK20B20
SK00B05
SK00B10
SK00B15
SK00B20
SK00B25
SK00B30
SK00B40

Soil composition
Sand
Silt
Clay
%
100
0
0
96
2
2
92
6
2
88
9
3
83
14
3
78
18
4
75
21
4
71
24
5
91
7
2
83
14
3
84
12
4
80
15
5
80
15
5
75
19
6
72
21
7
96
2
2
91
7
2
86
11
3
81
15
4
78
15
7
73
20
7
69
21
10

Limiting Density

Gradation
Cc

Cu
2
2
3
10
16
20
23
32
6
25
24
23
22
48
84
2
2
8
32
46
88
430

1
1
1
5
7
5
3
1
3
12
11
10
10
10
2
1
1
3
11
14
7
4

D50
mm
0.500
0.500
0.470
0.450
0.420
0.410
0.400
0.350
0.480
0.460
0.440
0.420
0.400
0.380
0.360
0.500
0.450
0.400
0.380
0.350
0.320
0.300

Replacement-of-void stage occurs when the void
spaces are almost completely filled. Addition of fineparticles would displace the sand grains from eachother; therefore, the density is decreasing with theaddition of fines content. The limiting value of fines
content at the transition from these two phases isknown as the threshold fines content (f th ), which
normally is located within 20% to 40% of fines
content. The threshold fines content of sandbentonite and sand-fine mixtures are both 20% offines by weight while for sand-kaolin is 25% ofkaolin by weight.

The maximum and minimum density of the sandmatrix soils are plotted against percentages of fines
by weight in Fig. 3. The negative hyperbolic curve is
noticeable from the plot of limiting densities. During
the filling-of-void phase, the density of soil isincreasing with fines content. The fine particles isfilling the void spaces between sand grains,
increasing the overall density of soil. Theliquefaction resistance would be lower because thepermeability is low; the rate of pore pressure build
up due to the external loading is therefore faster.

Density (Mg/m3)

2.0
1.8
sand-kaolin (min)1.6

sand-kaolin (max)1.4

sand-fines (min)sand-fines (max)

1.2
fth = 20 %1.00

sand-bentonite (min)fth = 25 %

10
20
30
Fines Content, FC (%)

sand-bentonite (max)40

Fig. 3 Limiting densities of sand matrix soils
274

GEOMATE- Bangkok, Nov. 14-16, 2016

CONCLUSION[3] Finn, W. D. L.. Assessment of liquefaction
potential and post-liquefaction behavior of earth
structures: developments 1981-1991. In Second
International Conference on Recent Advancesin Geotechnical Earthquake Engineering andSoil Dynamics, 1991, 1833–1850 .
[4] Prakash, S., and Puri, V. K. Recent advances inliquefaction of fine grained soils. In 5thinternational conference on recent advances ingeotechnical earthquake engineering and soil
dynamics, San Diego, California, 2010.
[5] Kramer, S. L. Geotechnical earthquake
engineering (1st ed) New Jersey: Prentice Hall,
1996.
[6] Tan, C. S., Marto, A., Leong, T. K., and Teng, L.
S.. The role of fines in liquefaction
susceptibility of sand matrix soils. Electronic
Journal of Geotechnical Engineering, 2013, 18
L, pp. 2355–2368.
[7] Tan, C. S., Marto, A., Makhtar, A. M., andPakir, F. Plasticity Behaviour of Sand Matrix
Soils. Applied Mechanics and Materials, 2015,
695, pp. 734–737.
[8] British Standards Institution. BS1377-2:1999
Classification tests. (2010a).
[9] British Standards Institution. BS1377-4:1999
Compaction-related tests, 2010b.
[10] Lade, P. V. and Yamamuro, J. A. Effects of
nonplastic fines on static liquefaction of sands.
Canadian Geotechnical Journal, 1997, 34(6), pp.
918–928.
[11] British Standards Institution. BS5930:1999
Code of practice for site investigations, 2010c.
[12] American Society for Testing and Materials.
D2487-11: Standard practice for classificationof soils for engineering purposes (Unified Soil
Classification System), 2011.

In this study, both primary and secondary
physical characteristics of sand matrix soils were
investigated through laboratory testing. The primary
physical characteristic including particle density,
particle size distribution, limiting densities were
examined. Based on the results, the following
conclusions are achieved:
1. The particle density of the clean sand,white kaolin and green bentonite used inthis study are 2.63, 2.62 and 2.35
respectively.
2. The threshold fines content of sandbentonite and sand-fine mixtures are 20%
of fines by weight while for sand-kaolin is
25% of kaolin by weight.
3. The physical characteristics are varying
with the amount of fines present within thematrix of sand grain.
4. Typical engineering characteristic of clean
sand could not represent as to all types ofsand matrix soils.
ACKNOWLEDGEMENTS
The authors gratefully acknowledged thefinancial supports given by the Ministry of
Education (MOE) under the Fundamental Research
Grant Scheme in undertaking this research.
REFERENCES[1] Wang, W. S. Some Findings in Soil
Liquefaction. Chinese Jounal of GeotechnicalEngineering, 1980.
[2] Seed, H. B., Idriss, I. M., & Arango, I..
Evaluation of liquefaction potential using field
performance data. Journal of GeotechnicalEngineering, 1983, 109(3), pp. 458–482.

275

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EFFECT OF FLY ASH ON THE STRENGTH DEVELOPMENT OFBANGKOK SOIL-CEMENT

Keeratikan Piriyakul1
Center of Excellence in Structural Dynamics and Urban Management, Department of Civil andEnvironmental Engineering Technology, College of Industrial Technology,
King Mongkut’s University of Technology North Bangkok, Thailand

1

ABSTRACTThis research presents the effect of fly ash on the strength development of Bangkok soil-cement mixing. The
Bangkok clay samples were prepared with 20 % of the Portland cement type 1 by weight of dry soil and varied
the replacement of fly ash type F 10, 15, 20, 25 and 30 % by weight of cement. Then, these samples were cured
to 7, 14, 28 and 90 days respectively and the bender element and the unconfined compression tests were
performed. The research found that the optimum percent replacement of fly ash was about 15-20 %. The Vs
result of 20 % replacement of fly ash at 90 days was the maximum Vs result, showing the long term behaviour of
fly ash mixed with cement. From the saturation equation, the Vs(max), the KVs, the G0(max) and the KGo parameters
showed that the optimum percent replacement of fly ash was about 15-20 %. In a similar way, the Su(max) and KSu
parameters showed that the optimum percent replacement of fly ash was about 15-25 %.
Keywords: Bangkok Clay, Fly-ash, Bender Element Test, Soil-cement
shear modulus, G0, in MPa is calculated using therelationship of elastic continuum mechanics asshown in Eq. (2).

INTRODUCTION
The shear wave velocity measurement technique
was used to determine the initial shear modulus, G0,
by means of elastic shear wave. This measurement
technique used the principles of wave propagation
showing a direct correlation between the shear wave
velocity and G0 as described in [1]. Piriyakul [2]used similar piezoelectric ceramic sensors to assessthe disturbance of soil samples. Piriyakul [3] usedpiezoelectric ceramic sensors to examine theanisotropy of clay soils by shear wave velocity.
Piriyakul and Gi-nga [4] and Piriyakul andPochalard [5] used piezoelectric ceramic sensors todetermine the cementation of soft Bangkok clay
mixed with cement and with fly ash, respectively.
This initial shear modulus is widely considered to be
an important parameter in earthquake engineeringand the prediction of soil-structure interaction. The
shear wave is generated and received by
piezoelectric transducers placed at opposite ends ofthe soil specimen. The shear wave velocity is
calculated from the tip to tip distance between the
two transducers and the time required by the shear
wave to cover this distance as shown in Eq. (1).
Vs = L/t

G0 = .Vs²(2)

where  is the mass density of the material in kg/m3.
The saturation equation, the so called MichaelisMenten equation [6], was used to predict thestrength development of Bangkok soil-cement asshown in Eqs. (3) to (5).
(3)
where Vs(max) is the maximum shear wave velocity
and KVs is the constant for shear wave velocity.
(4)
where G0(max) is the maximum shear modulus and
KGo is the constant for shear modulus.
(5)

(1)

where Su(max) is the maximum undrained shear
strength and KSu is the constant for undrained shear
strength.

where Vs is the shear wave velocity in m/s, L is thetip to tip distance between two transducers in mm,
and t is the required time to cover this distance in µs.
After obtaining the shear wave velocity, the initial
276

GEOMATE- Bangkok, Nov. 14-16, 2016that the two outer electrode surfaces are the samepole and the center electrode is the other pole. To
attach an electrical wire lead to the center electrode,a portion of the element must be ground away. The
series connected piezoelectric transducer is better to
use as receiver. On the other hand, the parallel
connected piezoelectric transducer is better to use astransmitter. However, this research uses only theparallel type for both transmitter and receivertransducers due to the advantage in measurement ofsending signal.

BENDER ELEMENT TEST

Fig. 1

Operation of piezoelectric ceramic sensor[1].

Bender element sensors were piezoelectric
sensors. The principle of piezoelectric sensor isbased on the properties of piezoelectric ceramicmaterials. A voltage applied to faces of a
combination of two piezoelectric ceramic materials
causes one to expand while the other contracts,
causing the entire element to bend as described by[7] and shown in Figure 1. Similarly, a lateral
disturbance of the piezoelectric sensor will produce
a voltage so the piezoelectric sensor can be used asboth shear wave transmitter and receiver.
Measurement of time delay between sending andreceiving of the shear wave will provide the shear
wave velocity.

Fig. 3

Schematic of the piezoelectric ceramic
sensor.

Figure 3 shows the piezoelectric transducer usedin this research. This sensor is a non-magnetic
piezoceramic with non-magnetic electrodes and nonmagnetic reinforcing materials. This sending sensor(T220-A4NM-303Y) is manufactured from the
Piezo System, Inc. The size of the sending sensor is
12.7 mm in width, 15.9 mm in length and 0.51 mm
in thickness. The research uses this sensor to send
the shear wave because of the strong sending signal.
Oscilloscope

Amplifier

Ch1 Ch2Computer
Soil sample
SendingsensorFig. 2

Receiving
sensor

Instrument
amplifierFig. 4 Schematic of shear wave measurement andassociated electronics.

a) Series and b) Parallel connected elements[7].

Figure 4 shows the schematic test set-up. A
personal computer generates a signal through asound card with 5V peak to peak as suggested by[8]. This signal is amplified to 40V peak to peak. An
oscilloscope is used to measure the arrival time
between a sending signal and a receiving signal. A
voltage pulse is applied to the sending sensor. This
causes it to produce a shear wave. When the shear
wave reaches the other end of the soil sample,
distortion of the receiving sensor produces anothervoltage pulse. The receiving sensor is directly
connected to the oscilloscope to compare thedifference in time between the sending and thereceiving signals. The shear wave velocity

There are two types of piezoelectric transducers.
One is a series connected piezoelectric transducer
and the other is a parallel connected piezoelectrictransducer. The series connected piezoelectrictransducer is shown in Fig. 2a. Noting that thepolarization is oriented in opposite directions foreach plate. An electrical wire lead is attached to eachof the outer electrode surfaces. The parallel
connected piezoelectric transducer is shown in Fig.
2b. In this second type of piezoelectric transducer,
the polarization has the same direction for bothplates. The electrical connections are attached such
277

GEOMATE- Bangkok, Nov. 14-16, 2016

measurements are usually performed withfrequencies ranging between 2 to 12 kHz, at strains
estimated to be less than 0.0001 %. At low
frequencies, signals can be influenced by a nearfield effect. At high frequencies, the receiving signal
is very weak and difficult to interpret. In most cases,
signals are averaged 32 times in order to get a clear
signal. The measurement of shear wave velocity insoil sample by means of piezoelectric ceramic
sensors is clearly described by [9] and [10].

fly ash. From the experimental results, the researchfound that the optimum percent replacement of fly
ash was about 15 %. From the 90 days results, the Suof 15 % replacement of fly ash was the maximum
value which reports the long term behaviour of fly
ash and cement mixed.
600

7 days550

14 days500

BANGKOK CLAY MATERIAL

28 daysVs (m/s)

450

The geological condition of Bangkok is reportedby [11] and shows that Bangkok is situated on alarge plain underlain by alluvial and deltaic
sediments of the Chaophraya basin. This plain is
about 13,800 km2 in area and is generally known as“the lower central plain”. The plain was under ashallow sea 3,000 to 5,000 years ago and theregression of the sea took place around 2,700 years
ago, leaving the soft deposits, which form the lower
central plain. This plain consists of thick clay known
as Bangkok clay on its top layer, and its thickness isabout 15 to 20 m in the Bangkok city area. The soft
clay has very low shear strength, and is highly
compressible, as it has never been subjected tomechanical consolidation.
The soft Bangkok clay at 10-12 m depth issampled for this research. The engineering
properties are found where that the specific gravity
is 2.72, the natural water content is 66.3%, the massdensity is 1,470 kg/m3 and the liquid limit is 88%.
60 experiment soft Bangkok clay samples areprepared at the liquid limit by mixing 20% by
weight of Portland cement Type 1 and varying the
replacement fly ash of 0, 10, 15, 20, 25 and 30%.
This research uses the fly ash Type F from Mae Moh
Lignite Power Plant in Lampang. Then, thesesamples are cured 7, 14, 28 and 90 days in order toperform the bender element test and the unconfinedcompression test.

90 days

400
350
300
250
200

0

10

20

30

40

% replacement

Fig. 5

Vs versus the % replacement of fly ash.
500
7 days450

14 days400

28 daysG0 (MPa)350

90 days

300
250

200
150
100

50
0

10

20

30

40

% replacement

Fig. 6G0 versus the % replacement of fly ash.
800

TEST RESULTS

7 days14 days
600

Figure 5 shows the shear wave velocity, Vs,
versus the percent replacement of fly ash. From thetest results, the research found that the optimum
percent replacement of fly ash was about 15-20 %.
From the 90 days results, the Vs of 20 %replacement of fly ash was the maximum value
which reports the long term behaviour of fly ash andcement mixed.
In a same way, Fig. 6 shows the shear modulus,
G0, versus the percent replacement of fly ash. Fromthe experimental results, the research found that the
optimum percent replacement of fly ash was 20 %.
In a similar ways, Fig. 7 shows the undrained
shear strength, Su, versus the percent replacement of

28 daysSu (kPa)

90 days
400

200

0
0

10

20

30

40

% replacement

Fig. 7 Su versus the % replacement of fly ash.
Figure 8 shows the plot of the 1/Vs versus the 1/t
278

GEOMATE- Bangkok, Nov. 14-16, 2016

0.010

varied with the percent replacement of fly ash. The
control Bangkok soil-cement sample data with 0 %replacement of fly ash depicted a Y-intercept of
0.00256. The invert of the Y-intercept value was theVs(max) of 390.63 m/s and the KVs of 2.36 wascalculated by using the Vs(max) of 390.63 m/s
multiplied with the slope of 0.00603. Table 1 shows
all results of Vs(max) and KVs parameters varied withthe percent replacement of fly ash. From the test
results, the research found that the optimum percent
replacement of fly ash was about 15-20 %. The
maximum value of shear wave velocity Vs(max) of 20% replacement of fly ash was 452.49 m/s and theKVs was 5.43 day for the Bangkok soil-cement todevelop strength half of Vs(max).
In a similar way, Fig. 9 shows the plot of the
1/G0 versus the 1/t varied with the percent
replacement of fly ash. Table 2 showed all results of
G0(max) and KGo parameters varied with the percent
replacement of fly ash.
In a same way, Fig. 10 shows the plot of the 1/Su
versus the 1/t varied with the percent replacement of
fly ash. Table 3 shows all results of Su(max) and KSu
parameters varied with the percent replacement of
fly ash.
0.0045

0%

10%

15%

20%

25%

30%

Linear (0%)

Linear (10%)

Linear (15%)

Linear (20%)

Linear (25%)

Linear (30%)

1/Su (kPa-1)

0.008

1/Vs (s/m)

0

0.08

0%
20%
Linear (0%)
Linear (20%)

0.012

10%
25%
Linear (10%)
Linear (25%)

20%: y = 0.02937x + 0.00225
R² = 0.93334
25%: y = 0.03359x + 0.00175
R² = 0.98734
30%: y = 0.03184x + 0.00318
R² = 0.91741% Fly ash
0
10
15
20
25
30

% Fly ash
0
10
15
20
25
30

0.16

15%
30%
Linear (15%)
Linear (30%)

0.006

0.002

0
10
15
20
25
30

20%: y = 0.04802x + 0.00311
R² = 0.84358
25%: y = 0.01309x + 0.00540
R² = 0.41789
30%: y = 0.02252x + 0.00645
R² = 0.34826

0.000
0.00

Fig. 90.05

0.10
1/t (day-1)

0.15

0.20

0.25

Vs (max)[m/s]
390.63349.65440.53452.49352.11327.87

K Vs
[day]
2.36
1.31
5.63
5.43
1.25
1.79

G0 (max)[MPa]235.29184.50317.46321.54185.19155.04

K Go
[day]
6.22
2.51
17.20
15.44
2.42
3.49

Table 3 Results of Su(max) and KSu parameters variedwith the percent replacement of fly ash% Fly ash

0%: y = 0.02642x + 0.00425
R² = 0.99221
10%: y = 0.01363x + 0.00542
R² = 0.50573
15%: y = 0.05417x + 0.00315
R² = 0.92456

0.10
0.15
1/t (day -1)

Table 2 Results of G0(max) and KGo parameters
varied with the percent replacement of fly
ash

0.008

0.004

0.05

Table 1 Results of Vs(max) and KVs parameters variedwith the percent replacement of fly ash

0.010
Go (MPa-1)

15%: y = 0.02202x + 0.00170
R² = 0.93250

0.004

Fig. 10 The relationship between 1/Su and 1/t.

The relationship between 1/Vs and 1/t.

0.014

10%: y = 0.03358x + 0.00316
R² = 0.86967

0.006

0.00

1/t (day -1)

Fig. 815%30%
Linear (15%)
Linear (30%)
0%: y = 0.01447x + 0.00184
R² = 0.81049

0.000

0%: y = 0.00603x + 0.00256
R² = 0.99778
10%: y = 0.00375x + 0.00286
R² = 0.52121
15%: y = 0.01279x + 0.00227
R² = 0.88209
20%: y = 0.01200x + 0.00221
R² = 0.74301
25%: y = 0.00354x + 0.00284
R² = 0.41053
30%: y = 0.00547x + 0.00305
R² = 0.36689

0.001510%
25%
Linear (10%)
Linear (25%)

0.002

0.00350.00250%
20%
Linear (0%)
Linear (20%)

0.20

The relationship between 1/G0 and 1/t.

279

Su (max)[KPa]
543.48316.46588.24444.44571.43314.47

K Su
[day]
7.86
10.63
12.95
13.05
19.19
10.01

GEOMATE- Bangkok, Nov. 14-16, 2016[3] Piriyakul K, “Examination of anisotropy of clay
using shear wave velocity”, The Journal ofIndustrial Technology, Vol.6, 2010, pp. 25-29.
[4] Piriyakul K & Gi-nga T, “Using shear wave
velocity to determine the cementation of soft
Bangkok clay mixed with cement”, Proceedingsof International Symposium on Sustainable
Geosynthetics and Green Technology for
Climate Change (SGCC), 20-21 June 2012,
Bangkok, Thailand, pp. 339-342.
[5] Piriyakul K & Pochalard, S, “Using shear wave
velocity to determine the cementation effect ofsoft Bangkok clay mixed with cement and flyash”, Multiphysical Testing on Soil and Shales,
Springer Series in Geomechanics andGeoengineering, 2012, pp. 311-316.
[6] Michaelis L & Menten M, “Die kinetik der
invertinwirkung”, Biochemistry Zeitung, Vol.
49, 1913, pp. 333-369.
[7] Kramer
SL,“Geotechnical
Earthquake
Engineering. Prentice-Hall International Series”,
Prentice Hall, Upper Saddle, New Jersey, 1996.
[8] Mohsin AKM & Airey DW, “Automating Gmax
measurement in triaxial tests”, The Prefailure
Deformation Characteristics of Geomaterials,
2003, pp. 73-80.
[9] Brignoli EGM, Gotti M & Stokoe KH,“Measurement of shear waves in laboratory
specimens by means of piezoelectrictransducers”, Geotechnical Testing Journal,
Vol. 19, 1996, pp. 384-397.
[10] Piriyakul K, “A development of a benderelement apparatus”, Journal of King Mongkut’s
University of Technology North Bangkok, Vol.
20, 2010, pp. 363-369 (in Thai).
[11] Tuladhar R, Yamazaki F, Warnitchai P & Saita
J, “Seismic microzonation of great Bangkok
area using microtremor observations”, J.of
Earthquake Eng Struct Dynam, Vol. 33, 2004,
pp. 211-225.

CONCLUSION
From the test results, the research found theconclusions as follows: The optimum percent replacement of fly ash
was about 15-20 %. From the 90 days results,
the Vs of 20 % replacement of fly ash was
the maximum value which reports the longterm behaviour of fly ash and cement mixed.
 From the saturation equation, the Vs(max) andthe KVs parameters showed that the optimum
percent replacement of fly ash was about 1520 %.
 From the saturation equation, the G0(max) andthe KGo parameters showed that the optimum
percent replacement of fly ash was about 1520 %.
 From the saturation equation, the Su(max) andthe KSu parameters showed that the optimum
percent replacement of fly ash was about
15-25 %.
ACKNOWLEDGEMENTS
The author is grateful to the scientific researchfund of the College of Industrial Technology and theKing Mongkut’s University of Technology North
Bangkok (KMUTNB).
REFERENCES[1] Dyvik R & Madshus C, “Laboratory
measurement of Gmax using bender elements”,
The ASCE Annual Convention, 1985, pp. 186196.
[2] Piriyakul K, “Soil disturbance assessment insoil sampling of open tube sampler and rotarycore drilling”, The Journal of Industrial
Technology, Vol.6, 2010, pp. 1-6.

280

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DRAINED SHEAR STRENGTH OF COMPACTED KHON KAEN
LOESS
1,2

R. Nuntasarn1 and W. Wannakul2
Faculty of Engineering, Khon Kaen University, Thailand

ABSTRACTThe objective of this study was to determine the relationship between drained shear strength and matricsuction of compacted Khon Kaen loess. Khon Kaen loess in this study was compacted at 95% by modified
method. This soil sample was investigated a soil water characteristic curve (SWCC) from the pressure plate test.
Moreover the relationship between drained shear strength and matric suction was determined from theconsolidated drained triaxial test (CD Test) under saturated and unsaturated condition. Three saturated
specimens were tested to determine an effective friction angle (f´) and an effective cohesion (c´). Another threeunsaturated specimens were tested with various matric suction of 30, 180 and 280 kPa, respectively. The net
confining pressures of these specimens were constant at 100 kPa. To determined drained shear strength atunsaturated condition, an effective friction angle (f´) was assumed to be constant with a matric suction value.
The SWCC of compacted Khon Kaen loess showed a bimodal curve. The first air-entry value and the firstresidual volumetric moisture content was 4 kPa and 20%, respectively. The results of saturated consolidated
drained triaxial test (CD Test) presented that an effective friction angle (f´) and an effective cohesion (c´) was 31
degree and 54 kPa, respectively. Moreover the unsaturated triaxial test showed a linear relationship between
drained shear strength and matric suction with the fb angle of 28 degree.
Keywords: Unsaturated Soil, Compacted Soil, Khon Kaen Loess, Consolidated Drained Triaxial Test, Soil
Water Characteristic Curve

degree before air entry value. After Air entry value
the relationship between cohesion intercept (c) andmatric suction was non-linear as shown in Figure 1.

INTRODUCTION
The unsaturated soil has three phases which aresolid, liquid and air. Therefore, the Terzaghi’s
effective stress law for saturated soil is not
appropriate for unsaturated soil. Reference [1]formulated the shear strength equation for an
unsaturated soil as given in Eq. (1).

tff = c′ + (σ n − u a ) tan f′ + (u a − u w ) tan fb (1)Where t ff is a shear stress at failure. c' is aneffective apparent cohesion, which is the shear
strength intercept when the effective stress is equal
to zero. σ f is total normal stress at failure. u af ispore-air pressure at failure. u wf is pore-waterpressure at failure. f' is an effective angle ofinternal friction. And fb is an angle indicating therate of increase in shear strength relative to the soil
suction at failure.
Reference [2] was investigated the shear strengthparameters of unsaturated residual soils by
consolidated drained test. This study found that thevalues of fb are generally lower than f′.
Reference [3] was studied consolidated drained
test on a residual soil, which was classified as CL
according to USCS. This study found that theeffective friction angle of 31.5 degree and a fb of 29

Fig. 1 the relationship between cohesion intercept (c)and matric suction (Ref [2])BASIC PROPERTIESThe soil used in this study was Khon Kaen loess,which was classified as silty sand (SM) according to
USCS. The basic properties of Khon Kaen loesswere present in Table1. Percent passing sieve No. 4
and No.200 of this soil sample was 100% and 47%,
respectively as shown in Fig.2. Moreover, a liquid281

GEOMATE-Bangkok, Nov. 14-16, 2016

The test result of a pressure plate and isopiestic
humidity showed the soil water characteristic curve(SWCC) as illustrated in figure 3. The soil sample
was compacted at the dry side of optimum moisture
content (6.5% or 95% of OMC). The SWCC ofcompacted Khon Kaen loess showed a bimodal
curve, which indicated two distinct air-entry values
and two distinct residual points [6]. The first andsecond air entry value is 8 and 5,000 kPa,
respectively. Moreover, the first and second residual
volumetric moisture content are 10 and 2%,
respectively. Therefore, the soil suction value ofsaturation regime and transition regime was between
0 to 8 kPa and 8 to 200 kPa, respectively. The soil
suction, which was higher than 200 kPa, was a
residual regime.
The cause of bimodal curve might be thespecimen was compacted at the dry side. Because
the SWCC of the undisturbed sample showed a
unimodal curve with two bending points as
illustrated in Fig 4.

limit and a plastic limit of this soil were 13% and of
12%, respectively, which gave the plastic index of
1%. The specific gravity was 2.65.
The maximum dry density and optimum
moisture content were 2.06 t/m3 and 8.5%,
respectively, which was determined by standard
compaction effort [4]. However, the specimen in thisstudy was prepared at 95% of maximum dry density,
which was 1.95 t/m3, and moisture content of 6.5%.
Table 1 Basic Properties of Khon Kaen loess
Properties
Liquid limit (LL), %Plastic limit (PL), %Plasticity index (PI), %Specific gravity
Optimum moisture content (OMC), %Maximum dry density (ρ d ), t/m3
Sand (% )Silt (%)
Clay (%)
USCS classification
120

Silt

Sand

110

13
12
1
2.65
8.5
2.06
53
27
20
SM
Clay

100
90
80

%Finer

70
60
50
40
30
20
10
0
10

1

0.1
Particle diameter (mm)0.01

Fig. 3 Soil water characteristic curve of compacted
Khon Kaen loess

0.001

Fig.2 Grain size distribution
Volumetric moisture content (%)

35

SOIL WATER CHARACTERISTIC CURVEThe drying soil water characteristics curve(SWCC) for the compacted Khon Kaen loess wasdetermined by two methods. According to [3], thepressure plate method was used to establish SWCC
for a suction values between 1 to 1,500 kPa.
Moreover, SWCC at suctions above 1,500 kPa were
determined from the isopiestic humidity method. In
this method, three solutions, which were Copper
Sulphate (CuSO4), Ammonium Chloride (NH4Cl)
and Sodium Hydroxide (NaOH.H2O), were used todetermine SWCC at a suction value of 3,900 kPa,
30,900 kPa, and 365,183 kPa, respectively. Data
points above 1,500 kPa were total suction values.

30
25
20
15
10
5
0
0.1

1

10

100

1000

10000

100000 1000000

Soil Suction (kPa)

Fig.4 Soil water characteristic curve of undisturbed
Khon Kaen loess282

GEOMATE-Bangkok, Nov. 14-16, 2016

shear stress (t max ) as shown in Eq.(2) and Eq.(3),respectively. Then the cohesion intercepted (c") was
determined from Eq.(4). And a fb angle wasdetermined from Eq.(5).

CONSOLIDATED DRAINED TEST
Soil sample was mixed with water to achieve
1.95 t/m3 in dry density and 6.5% of moisture
content. Then soil sample was equally separated in
two parts and statically compacted in a molddiameter of 50 mm and 100 mm of height.
Each samples were saturated by applied
backpressure, which was lower than confiningpressure (i.e. σ′c = 20 kPa) prior to consolidate anddrained shear. B-values were observed to be greaterthan 98% to complete saturation.
Three net confining pressures of 100, 200 and
300 kPa were investigated in this study. For thesaturated series, three specimens were tested todetermine the effective friction angle (f′) andcohesion (c′). Pore air pressure (u a ) and pore waterpressure (u w ) of these three samples were controlled
at 20 kPa. For the unsaturated series, three samples
were observed at net confining pressures of 100 kPa.
Pore air pressures were controlled at 50, 200 and
300 kPa and pore water pressures were controlled at
20 kPa to achieve a matric suction of 30, 180 and
280 kPa. The shear rate in this study for bothsaturated and unsaturated series was 0.01 mm per
minute.
The failure of both saturated and unsaturated
sample showed a failure plane of 60 degree asshown in Fig.5. For both saturated and unsaturated
series, the stress-strain relationship showed the strain
hardening behavior as illustrated in Fig.6. Thus thefailure was defined at 20% strain. The test result of
saturated sample was present at Table 2. Then the
Mohr’s Circle was drawn in Fig.7. The Mohr’s
circle showed that the effective friction angle andcohesion was found as 31 degree and 44 kPa,
respectively.

(a) Saturated series(b) Unsaturated series
Fig. 6 The relationship between stress and strain ofcompacted Khon Kaen loess
Table 2 A saturated sample result

600(a) Saturated series

σ3

ua

u wf

σd

t max

(kPa)

(kPa)

(kPa)

(kPa)

(kPa)

120

20

20

350

174

220

20

20

565

282

320

20

20

799

399

τφφ = τ max . cos φ′(b) Unsaturated series

σneτφ =

Fig.5 A shear failure of compacted Khon Kaen loess(σ1neτφ + σ3neτφ ) − τ

2c′′ = τφφ − σ neτφ . τan φ′

The drained shear strength of unsaturated soil
was calculated by assumed constant f′ (f′ =31 degree) as present in Fig.8. The shear stress (t ff )and the net stress (σ netf ) at failure plane ofunsaturated soil were calculated from the maximum

 c′′ − c′ 
φb = tan −1  (u a − u w )φ 

283

(2)

max . σin φ′

(3)
(4)
(5)

GEOMATE-Bangkok, Nov. 14-16, 2016

500

200

300

c'' (kPa)t, kPaTransitionl
Regime250

f' = 31o
400

200
100
44

150

Residual
Regime

100
50

0
0

100

200

300

400

500

600 700σ´n, kPa

800

9001000 1100 1200

00

Fig.7 The Mohr’s circle of compacted Khon Kaenloess at saturated condition

50

100
150
200
Soil suction (kPa)

250

300

Fig. 9 The relationship between drained shear
strength and matric suctiontf'
σnet, tmax

σnetf, tfff'c′′
σ3netfσnet

σ1netfFig.8 The interpretation of unsaturated drained shear
strength

The test results of unsaturated sample were
present in Table 3. The plot between drained shear
strength and matric suction was illustrated in Fig.9.
Figure 9 showed that the drained shear strength of
this study was investigated at saturated, transition
and residual regime. In addition, Fig.9 presented a
non-linear relationship between cohesion intercept
and matric suction. Table 3 showed that fb wasslightly lower than f′. Moreover, fb was constant at
30.5 degree for the matric suction range between 0
and 180 (fb was constant before residual regime).
Then matric suction was higher than 180 kPa, a fb
angle value was dropped from 30.5 to 28 (fb wasdecreased after the transition regime).
Moreover, the relationship between the total
volumetric strain and axial strain showed the dilation
behavior of both saturated and unsaturated series as
illustrated in Fig.10 and Fig.11, respectively. The
dilation behavior is the characteristic of dense soil.
The relationship between the water volume
change and axial strain of unsaturated sample aspresent in Fig.12 showed that water was suction intoa specimen.

Fig. 10 The relationship between the total
volumetric strain and axial strain of saturated sample

Fig.11 The relationship between the total volumetricstrain and axial strain of unsaturated sample

CONCLUSIONSThe soil sample in this study was Khon Kaen loess,which was classified as silty sand. This study foundthat the SWCC of compacted Khon Kaen loess wasa bimodal curve. The cause of bimodal curve might
be the specimen was compacted at the dry side. The
first and second air entry value is 8 and 5,000 kPa,
respectively. Moreover the first and second residual
volumetric moisture content is 10 and 2 %,respectively. The stress-strain behavior of

Table 3 An unsaturated sample results
σ3

ua

u wf

σd

t max

t ff

σ netf

c′′

(kPa)

(kPa)

(kPa)

(kPa)

(kPa)

(kPa)

(kPa)

(kPa)

fb
()

150
300
400

50
200
300

20
20
20

437
759
896

219
379
448

187
325
383

206
283
317

63
154
193

30.5
30.5
28.0

o

284

GEOMATE-Bangkok, Nov. 14-16, 2016

REFERENCES

compacted Khon Kaen loess for both saturated andunsaturated series showed a strain hardening. In
addition the relationship between the volumetricstrain and axial strain of both saturated andunsaturated series showed the dilation behavior. The
consolidated drained triaxial test of saturated
compacted Khon Kaen loess presented the effective
angle and cohesion of 31 degree and 44 kPa,
respectively. The unsaturated drained shear strength
was determined by assumed a constant friction angle(f′). fb was constant before residual regime. Afterthe transition regime, fb was decreased.

[1] Fredlund, D.G. and Rahardjo, H. Soil Mechanics
for Unsaturated Soils. Wiley. New York. 1993
[2] Huat, Bujang BK, Faisal Hj Ali, Affendi
Abdullah, Saravanan Mariappan, and Soenita
Hashim. Shear Strength Parameters of
Unsaturated Residual Soils. Jurnal Kejuruteraan
17. 2005[3] Rahardjo, H., Ong, B.H. and Leong, E.C. Shear
Strength of Compacted Residual Soil From
Consolidated Drained and Constant Water
Content Traxail Test. Canadian Geotechnical
Journal, 41: 421-436. 2004
[4] ASTM - American Society for Testing Material.
Standard Test Method for Laboratory
Compaction Characteristics of Soil using
Standard Effort (12,400 ft-lbf/ft3 (600 kNm/m3)) (ASTM D698-00a). Ann Arbor.
Michigan. 2000
[5] ASTM - American Society for Testing Material.
Standard Test Method for Capillary-Moisture
Relationships for Coarse- And Medium-Textured
Soils By Porous-Plate Apparatus (ASTM D232598). Ann Arbor. Michigan. 1998

Fig.12 The relationship between the water volume
change and axial strain of unsaturated sample
ACKNOWLEDGEMENTS[6] Gitirana Jr., Gilson de F.N. and Fredlund, D.G.

Acknowledgement is given to Faculty ofEngineering, Khon Kaen University for the supportof this re-search. Sincere thanks are extended to Dr.
Ketsuda Dejbhimon and Mrs Pornpis Chuson, forthe favor of using instruments in this study.

Soil –Water Characteristic Curve Equation with
Independent Properties. Journal of Geotechnicaland Geoenvironmental Engineering. 2004.Vol.
130. No.2. pp.209-211

285

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

A CASE STUDY ON SOIL IMPROVEMENT FOR REFORMING THE
BED OF A CYLINDRICAL STORAGE TANK USING MICROPILES.
Mohammad reza Atrchian1, Ali NisariTabrizi1
1

Department of Geotechnic Engineering, Faculty of civil engineering, Islamic Azad University, Zanjan, Iran.
ABSTRACT

Problem statement: Unsymmetrical immediate settlement is one of the defects from storage tanks located in the
bed composed of different subsurface soil. In this study, the tank was established in the bed of marl layers andcoral limestone layers with many holes. Moreover, gaps in the boundary both kinds of layers (marl layers andlimestone coral layers) have also been observed. These gaps could lead to increase in the contingency of an
unsymmetrical immediate settlement following loading of the tank. The aim of this study is to select a suitable
method of soil improvement for reforming the bed of a cylindrical storage tank. Approach/methodology: There
was a limitation on development of the options as the tank structure had already been established. Therefore, themost appropriate of remedial option, micropile along with injection under the pressurewas detected. Results: Theprocess of micropile along with injection under pressure leds to reduction of the soil settlement, as loads wereappropriately distributed to ground. In addition, with filling holes, without appearance of the unsymmetrical
immediate settlement, loads were transferred to the lower layers. Conclusions: Clearly, with the performance of
micropiles along with injection under pressure have also been improved the soil characteristics.
Keywords: Soil improvement, Unsymmetrical immediate settlement, Micropiles

is a high probability of primary consolidation after
the tank being loaded.
The proper method of soil improvement for thelocation of the tank has to be selected according to
special conditions of the site. In the situation when
the tank had not been built, there were more options
for the soil improvement plan. There will be fewer
options now that the tank has been built and thestructure is located on the specified area.
There are two important issues which have to besolved by the proper improvement method
design:1- Filling the pores present in the corallimestone layer and in the boundary
between the coral limestone and Marl.
2- Proper load transference to the lower layersof the ground.

INTRODUCTION
There are generally two methods in front of thegeotechnical engineer for dealing with problematicsoils such as loose soils with low bearing capacity,
soils with high consolidation capacity, liquefaction
potential, and loose fills:
1- Utilization of load bearing elements in soil
2- Soil improvement and modification of thephysical-mechanical properties of the soilmass[1], [2].
In this regard, one of the methods of soil
improvement is by using micropiles [3]. Micropiles
are used to improve the properties of the soil andreinforcement of the ground beneath a foundation[4]. In the current study, the soil underneath the site
in the vicinity of a tank, mainly consisting of corallimestone with a lot of pores on a bed of Marl is
being investigated. The site had to be leveledin
order to build the tank, and some parts of the groundhad to be filled and excavated subsequently. The
studied tank is located between the excavated andfilled ground parts. According to inspections andsite investigations, existence of large voids, and lotsof pores is evident in the coral limestone layer.
Furthermore, some openings have been detected in
the boundary between the Marl and coral limestonelayers. Due to the existence of these openings, there

According to the soil model underneath the tank(with the assumption of the pores being filled), andthe results of the analysis of the soil under the tankusing PLAXIS software, primary consolidations
were found to be greater than the allowed value.
As seen by the results, an appropriate method has
to be used to reduce the sum of primaryconsolidation and secondary compression in the soil
beneath the foundation of the tank.
In the current study, the selected solution for thetwo mentioned issues is by using micropiles alongwith grouting.

286

GEOMATE-Bangkok, Nov. 14-16, 2016

METHOD AND MATERIAL

Information obtained from the drilled boreholes havebeen used to calculate the volume of grouting
material, which might be subject to some variations(due to undetermined state of soil and rock layers
under the tank). A summary of the calculations isgiven in Table 2.

Construction Method
As specified by the existing reports and the stateof the tank due its location in the boundary region of
excavated and filled zones, there is a probability of
non-uniform consolidation underneath. Furthermore,
the pores which are distributed in an unspecificmanner have to be filled with the proper materials.
According to the suggested improvement plan which
consists of implementing micropiles with appropriate
grouting, PLAXIS 8.2 for two dimensional modelinghas been used to simulate the soil under the tanks andthe micropiles with grouting (Fig. 1). Mechanicalproperties of the soil from the boreholes of the site
have been used to model the soil under the tank. It has
to be noted that the properties of the improved corallimestone are used in the modeling process; and also
the micropilesare located at angles of 45, 70, and 85
degrees to the horizon around the tank.

Table 2
Tan Thicknes
ks of rock
area
layer
(M2)
(M)

No. 32 bars have been used for the micropiles.
Weight of the bar and the casing used for micropile
construction of the tanks is given in Table 3.
Table 3
Angel ofmicro piles tothe horizontal(Degree)45

In consistence with the design plan, 345
micropiles were drilled in the vicinity of the tank. The
amount of drilling operations for the execution of the
micropiles is listed in table 1.

45
70
85

Sectional
area

0.07
0.07
0.07

Number23

115

Weightof the
bar
(Kg)
16642

20
115
14523
20
115
14523
Sum total (Kg)
~ 46000
An estimate of the bars used for the micropiles
Micropile construction briefing

Volume(M3)

The following steps have been taken for theconstruction of the micropiles:• Preparation of the tools and utilities for theconstruction of the micropiles including
boring and grouting machines.
• Installation of the central grout mixing
station.
• Drilling the boreholes according totechnical specifications.
• Installation of permanent steel casing and
No. 32 bar in the micropile• Installation of spacers inside the casing for
grouting the micropiles.

(M2)

115
23
115
20
115
20
Sum total (M3)

Length of
the bar(M)

70
85

Table 1
Amount of drilling operations of the micropiles
Length(M)

Injection
volumeofmicropile
s
(M3)
*500

Weight of the bars

Drilling the micropiles

Number

Pores
volum
e
(M3)

466
2
25.4
2368
3
466
8
20
14921
3
Sum total (M3)~18000Volume of the grouting operations for tank.
*- Volume of the grouting has been calculated withthe reduction of the area of the bar and casing (ifused any) taken into account.

Properties of the modeled soil layers and thegrout surrounding the micropile are given in Figure 2.
Sum of the primary and secondary consolidation
under the tank is equal to 20.4 centimeters based on
the results from the program calculations.
Additionally, the sum of the acceptable minimum
primary and secondary consolidation, which canoccur under the tank in, is 9.5 centimetersfor the filledarea, and 6 centimeters for the excavated areasubsequently.

Angel ofmicro piles
to thehorizontal(Degree)

Porosit
y
(%)

185
161
161
~510

Grouting
Existing pores in the limestone layer and also thedrilled boreholes have to be filled with propermaterial based on the soil improvement design plan.
287

GEOMATE-Bangkok, Nov. 14-16, 2016••

on the grouting method in the soil. There are multiple
ways to construct a micropile but it always leads toinserting a bar or a tube with a few centimeters indiameter which is secured in the ground by means ofgrouting. Hence, construction method of micropiles
is such as the regular piles [5]. The main forces which
are generated in the micropiles are mainlycompression and tension [6]. Due to small diameterof micropiles, small force on the tip of the micropile
is assumed to be negligible and hence the interaction
between the soil and micropile is most frictional [6].
Piles are categorized into two main groups:
1. Displacement piles: piles which are driven
into the ground by means of vibration andpressing, which cause the surrounding soil tobe displaced [6].
2. Replacement piles: these piles are inserted
into pit holes previously dug, and replace thesoil within the hole [7].

Preparation of the grout mixture (water tocement ratio of around 0.35 to 0.6)
Grouting with maximum pressure of up to 5
atmospheres (0.5 KPa) of the micropiles.

RESULTS
After performing the soil improvement planunder the tank, excavated and filled zones were
locked together and all the existing pores have been
filled either with a mix of cement and stone dust
mortar or a mix of cement and soft sand. Micropilesof 300 millimeters in diameter, 23 meters in length,with 45 degrees to the horizon; and also some otherwith 70 and 85 degrees to the horizon and 20 meters
long have been implemented under the tank withcenter to center distances of 3 meters. Spacing of the
micropiles with 45, 70, and 85 degrees to the horizon
is equal to 1 meter as indicated in Figure 3. Micropile
spacing and angles have been determined based on
the significance of mortar penetration in the vicinityof the tank, and also its penetration extent under thetank. The various degrees of micropiles cause themortar to be distributed in a wider range.
It has to be noted that the length of the piles has
been designed to penetrate into the strong Marl layer
and transfer the load in a sufficient length. During theexecution of these piles, grouting was done to fill thepores and continued until they were filled completely.

In the current study, micropiles are built using
the second method (replacement method) by meansof drilling the hole.
Load beating capacity of the micropiles are often
dependent on the steel components with high load
bearing capacity inside, which are capable of bearing
most of the load.
Furthermore, in the current study, single
micropiles or a group of micropiles were used which
are loaded directly and the main part of the load is
supported by the reinforcement and can be used as analternative for regular piles for transferring the overload to lower layers of the soil.
Using micropiles has some advantages overusing other soil improvement methods which are:
1- Using micropiles causes the least amount ofdisturbance for nearby structures, soil, and theenvironment.
2- Micropiles can be constructed in every type ofsoil, landfill, rock, ground conditions and areas
with limited access.
3- Micropiles can be constructed with any angle inthe soil.
4- The equipment used to for micropile constructionis the same as the tools used for construction of
tiebacks, soil nail walls, and grouting projects.

After the execution of the micropiles, hydrostatic
test was performed according to segment 5-6-3-7 ofAPI 650 guidelines, which showed an improvementin the settlement conditions based on the results fromsettlement analysis of primary and secondary
consolidations,as well as the output from PLAXIS
software (Fig. 4 and 5). Therefore, execution of the
micropiles has caused the overload to be transferred
to stronger lower layers of soil, limit the settlement,
seismic improvement, reinforcement of foundationand the soil under the tank, and improvement of thearea soil. Mechanical properties of the soil in the
project site and around the micropiles have changed
using the micropiles, and mechanical parameters suchas compactness, load bearing capacity, permeability,
and compressibility modulus of the soil have been
improved.

CONCLUSION
In the current study, due to the limitations in thesoil improvement method imposed by existence of thetank, the best solution was recognized as utilizationof micropiles with grouting.Construction of
micropiles lead to the loads being transferred to thelayer with a sufficient bearing capacity and hence, a
reduction in settlement under the tank. Furthermore,pressurized grouting lead to the pores in the soil being
properly filled which helped the load be transferred to
the lower layers of the soil without creating any
unacceptable primary consolidation. Additionally,

DISCUSSION
Micropiles are used to improve the properties ofthe soil and reinforce the ground under a foundation[3]. Soil improvement is achieved by pushing asufficient number of micropiles and the resulting
effect is assumed to be the generation of strong
composite materials. Micropiles have diameters lessthan 300 millimeters and their performance depends
288

GEOMATE-Bangkok, Nov. 14-16, 2016

Soil properties also improved with construction of the
micropiles.

Fig.1. Simulation of the tank No.With PLAXIS

289

GEOMATE-Bangkok, Nov. 14-16, 2016

Fig. 2. Properties of the soil used in the modeling with PLAXIS

290

GEOMATE-Bangkok, Nov. 14-16, 2016

Fig. 3. Three dimensional view of the micropile execution under the tank.

Fig. 4. Primary consolidation under the tank with micropiles291

GEOMATE-Bangkok, Nov. 14-16, 2016

Fig.5. Secondary consolidation under the tank with micropiles
REFERENCES[1] Horn R, Taubner H, Wuttke M, Baumgartl T. Soilphysical properties related to soil structure. Soiland Tillage Research.30. 1994, pp.187-216.
[2] Meyerhof GG. Some recent research on the
bearing capacity of foundations. Canadian
Geotechnical Journal.1. 1963, pp.16-26.
[3] Babu SG, Murthy SB, Murthy D, Nataraj M.
Bearing capacity improvement using micropiles acase study. American Society of Civil Engineers.
proceedings of sessions of the geosupport
conference 2004, pp.1-8[4] Alnuaim A, El Naggar M, El Naggar H.
Performance of micropiled raft in clay subjected
to vertical concentrated load: centrifuge modeling.

Canadian Geotechnical Journal. 52. 2015,
pp.2017-2029.
[5] Juran I, Bruce D, Dimillio A, Benslimane A.
Micropiles: the state of practice. Part II: design ofsingle micropiles and groups and networks of
micropiles. Proceedings of the Institution of Civil
Engineers-Ground Improvement. 3. 1999, pp.89110.
[6] Han J, Ye S-L. A field study on the behavior of
micropiles in clay under compression or tension.
Canadian Geotechnical Journal.43. 2006, pp.1929.
[7] Nogami T, Novák M. Soil pile interaction invertical vibration. Earthquake Engineering &Structural Dynamics. 4. 1976; pp.277-293.

292

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

BEHAVIOR OF STEEL NAILS-REINFORCED STONE COLUMNMADE OF MIXTURE OF TYRE CHIPS AND STONE AGGREGATESTanwee Mazumder1, Nishant Neeraj2 and Ramanathan Ayothiraman3
Ph.D. Scholar, Former M.Tech Student, 3Associate Professor, Dept. of Civil Engg, IIT Delhi, New Delhi,
India

1

2

ABSTRACT
Scrap tyre generated from passenger cars, light and heavy trucks, heavy equipment, aircraft, and off–road
vehicles have many good engineering properties and hence has been effectively utilized for civil engineering
applications such as lightweight material for backfill of retaining structures, drainage layer and thermal
insulation layer etc. Typically stone aggregates of size 15 mm to 75 mm are used in the construction of stone
column. A few studies reported use of spent ballast and construction debris as a substitute to stone aggregates instone columns. In the present study an attempt has been made to investigate the behavior of stone columns
reinforced with vertical steel nails and constructed with mixture of stone aggregates and shredded tyre chips
through a series of small scale model experiments. Model stone column of diameter (d) 100 mm and l/d ratio of
4.5 is used to simulate floating stone column. The stone columns were reinforced with vertical steel rods to theentire length of column. Loading is applied through a hydraulic jack and reaction loading frame. Different mix
proportions of stone aggregates and tyre chips: 100S, 70S+30T, 50S+50T, 30S+70T and 100T were used inmodel testing. Results of model tests indicates that the conventional columns can be replaced by columns madeup entirely of shredded tyre chips when reinforced with vertical rods as encasement along the column
circumference.
Key Words: Tyre Chips; Stone Column; Encasement; Soft clay.

frictional strength) [19]. However, limited studies on
the application of shredded tyre shreds as aggregatesin ordinary stone columns have been carried out [2021]. Therefore, in this study vertical steel nails areused to encase the columns made of different mix
proportions of shredded tyre chips and stoneaggregates through model tests.

INTRODUCTION
Soft soil necessitates improvement of the groundso that the ground can be used beneficially forvarious projects that serve to mankind. Thus, torender it useful for various construction or suchother purposes, soft soil needs to bestabilized/reinforced so that it can bear the loads
coming from superstructure. Ground improvementby the installation of stone columns have been awidely accepted technique as it results in reduced
settlements and accelerates the rate of consolidation
settlement [1]. Various studies have been carried out
on the mechanism of failure of stone columns andthe load settlement behavior as well as on the designof ordinary and encased stone columns [2-17].
Also the disposal of scrap tyres produced in hugequantities every year across the globe poses hugeenvironmental threat. The amount of scrap tyres
produced every year is also increasing at an
alarming rate. Thus, this study looks into theviability of using scrap tyres as an alternative tostone aggregates in stone columns. It is seen in theliterature that apart from the stone aggregates usedin conventional stone columns, various otherrecycled materials like spent railway track ballast
and crushed concrete debris are being used as
aggregates in stone columns [18]. Tyre chips can bebeneficially used in geotechnical applications due toits advantageous properties (low density,hydrophobic nature, durability, resilience and high

MATERIALS USEDThe various materials used in the modelexperiments and their properties and characteristics
are presented below.
Soil
The clay bed in the model experiments wasconstructed of Kaolinite clay. Kaolinite clay isshown in fig. 1. The properties of clay is given in
Table 1.

Fig. 1. Kaolinite clay
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GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Properties of Kaolinite Clay
Property
Liquid limit
Plastic Limit% Clay% SiltOptimum moisture content
Maximum dry density
Shear strength at w = 38%

Encasement/Reinforcement
Vertical rods, having 8 mm diameter were
provided around the circumference of the stone
column to encase the stone column constructed at
the centre of the clay bed. The typical view of nailreinforced column in both plan and section is shownin fig. 4. The image of nails used is shown in fig. 5.

Value
45%
23%
62%
38%
21%
1620 kg/m3
7.5 kPa

Stone aggregates and shredded tyre chips
Stone aggregates and shredded tyre chips passing
thorough 12.5 mm sieve and retained on 10 mmsieve was taken to construct stone column in the test
tank. Laboratory tests were conducted to determine
the various index and engineering properties of thestone aggregates and tyre chips. Stone aggregate andshredded tyre chips are shown in figs. 2 and 3
respectively. The properties are listed in Table 2.

Fig. 4. Top view and cross-sectional view of nailreinforced column

Table-2. Properties of Tyre Chips and Stone
Aggregates
Property
Average Size
Specific Gravity
Maximum Density(kg/m3)
Minimum Density(kg/m3)

Tyre
Chips
10 mm
1.10

Stone
Aggregates
10 mm
2.78

575

1608

527

1367
Fig. 5. Vertical steel nails
Test Setup
A cylindrical tank of wall thickness 10 mmhaving a diameter 550 mm and height 750 mm wastaken to conduct model experiments. For the stone
columns, a scaling of 4 to 10 has been adopted as thecolumn diameter in the field varies between 400 to1000 mm. As for the steel rods a scaling factor of 4
is assumed, that is the 8 mm diameter used in themodel tests can be replaced by 32 mm rods in thefield. Also it is difficult to ensure complete verticalinstallation of steel rods of small diameters like 8
mm over great depths. Hence bigger size diameter
steel rods e.g. 32 mm rods can be installed to greater
depths. The loading arrangement involves loading
through a reaction loading frame.

Fig. 2. Stone aggregatesIn this study, tests were performed on 100 mm
diameter columns having l/d ratio of 4.5.
Replacement method was employed to construct thestone column at the center of the clay bed. A sandblanket of 30 mm thickness was placed over theequivalent circular loading area (2 times theFig. 3. Shredded tyre chips
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GEOMATE- Bangkok, Nov. 14-16, 2016

diameter of the stone column) to facilitate betterstress transfer. Above the sand blanket, a loading
cap having a diameter of 200 mm and 15mm
thickness was kept to apply the load centrally anduniformly. The load was applied through a hydraulicjack and a reaction frame. These arrangements areshown in fig. 6.

mm was maintained between the consecutive rods.
The clay was scooped out in stages using a helicalauger. Tyre chips/stone aggregates were charged
into the hole in layers of 50 mm and each layer was
given light uniform compaction. Casing pipe wasraised in stages ensuring minimum 10 - 20 mm
penetration below the tyre chip/aggregates placed.
Moisture Content (%)
30

40

50

60

Depth (cm)

0
10
20
30
40
50
60

100T

70

Fig. 7. Variation of moisture content with depth ofclay bed
Shear strength (kPa)

1-Loading frame, 2-Proving Ring, 3-Hydraulic jack,
4-Reference frame for fixing the Dial gauge, 5-Sand
Mat, 6-Cylindrical test tank with clay bed, 7-Stone
column, 8-Dial gauge.

0

1

2

3

4

5

6

7

8

9

10

0

Fig. 6. Test setup.

Depth (cm)1020

Clay bed preparation

30

The soft clay bed was made of remoulded soft
Kaolinite clay. The water content was maintained at
38 % to obtain very soft clay bed (c = 7 kPa) ofshear strength 7.5 kPa. Clay was uniformly mixed
with water until no lumps were present. The testtank was filled in layers with the paste and eachlayer was compacted by hand kneading. The
finished clay bed was covered with wet clothes andleft for 24 hours. The moisture content readings andshear strength were taken at every 100 mm depth ofclay layer. The typical variation of moisture andshear strength over the depth of clay bed for 100T
column is illustrated in figs. 7 and 8 respectively. It
is seen from figs. 7 and 8 that the water content andshear strength are uniform throughout the depth oftank, thus ensuring homogeneity of clay bed.

40
50
60

100T

70

Fig. 8. Variation of shear strength with depth of clay
bed.
Loading of stone column
A sand mat is placed over the column and it istopped with a loading cap of 15 mm thickness and
200 mm diameter. The load was applied through ahydraulic jack and pump system supported by areaction loading frame. The settlements wererecorded by a dial gauge. Load settlement and time
settlement curves were plotted for analyzing the
behavior of model the stone columns. The complete
arrangement is already shown in above Fig. 6.

Stone column construction
The stone column was constructed byreplacement method. The centre of the cylindricaltank was properly marked. Silicon grease wasapplied on the inner and outer surface of a thin openended seamless Perspex cylindrical pipe of outer
diameter 100 mm the pipe was pushed into thecentre of the clay bed upto a depth of 450 mm. Ironrods of diameter 8 mm were then pushed into theclay bed along the circumference of the perpex pipe
to install these rods vertically. A clear spacing of 27

DISCUSSION OF RESULTS
The model tests were performed on vertical mild
steel nails-reinforced stone column made of different
mix proportions of stone aggregates and shredded
tyre chips to study the time dependent load
settlement behavior. Similar load tests were also295

GEOMATE- Bangkok, Nov. 14-16, 2016carried on pure unreinforced clay bed and clay bed
without stone columns but provided with steel nails
circumferentially over a circular area locatedcentrally over the clay bed and having a diameter of
100 mm. The typical measured time-settlement
response for 100S, 50S+50T and 100T are shown in
figs. 9-11. Figures 9 - 11 demonstrate that thesettlement for each load increment is found tostabilize at around 70 to 100 mins, for various mix
proportions of stone aggregate and shredded tyrechips.

undergo lesser settlement, while 100T column
undergo higher settlement for the same load. This
indicates that as the percentage of tyre chips
increases, the columns settle relatively high.
Nevertheless, the load carried by the stone columns
is always higher than the load carried by only clay
bed and clay bed reinforced with nails only.
Load (N)
0

200

300

400

500

600

Settlement (mm)

0
25
50

Increment1
Increment2
Increment3
Increment4
Increment5
Increment6

75

Settlement (mm)

100

1000

1500

2000

0

Time (min)
0

500

2500

3000

100S
70S+30T
50S+50T
30S+70T
100T
clay rods
Pure clay

25

50

75

100

Fig. 12. Comparative study of load-settlement
behavior of different stone columns.

100

Fig. 9. Time-Settlement response of 100S column
Time (min)
0

100

200

300

400

500

600

Settlement (mm)025

The load carrying capacity of stone columns aretaken by double tangent method for stone columns
made of various mix proportions as well as that ofclay bed with iron rods and pure clay beds, and thevalues are presented in Table 3.
Table 3 Ultimate load carrying capacity of various
stone column.

50

Increment1
Increment2
Increment3
Increment4
Increment5
Increment675

Stone column
type
100S

Experimental capacity(N)

70S+30T

1952

50S+50T

1750

30S+70T

1492

100T
Clay bed with ironrods
Pure clay bed

1208

100

Fig. 10. Time-Settlement response of 50S+50T
columnTime (min)
0

100

200

300

400

500

600

Settlement (mm)

0

25

50

75

100

2471

413
313

From Table 3 it can be seen that encasing thecolumn with vertical nails cause a significant rise in
the ultimate load carrying capacity for each of themix proportion. The load carrying capacity ofordinary stone column made of 100S and 100T istheoretically calculated using IS code method (IS:15284, 2003) and the values are 1137 N and 653 N
respectively. It is seen that reinforced column madeof all mix proportions exhibits higher capacity thanordinary column. Thus, reinforcing the stone
columns increases the stiffness and reduces the

Increment1
Increment2
Increment3
Increment4
Increment5
Increment6

Fig. 11. Time-Settlement response of 100T column
The load-settlement curve for 100S, 50S+50T
and 100T is shown in fig. 12. It is seen from thefigure that the columns made of 100S and 50S+50T
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GEOMATE- Bangkok, Nov. 14-16, 2016

severe lateral bulging signiﬁcantly and hence
enables the encased columns to exhibit higher load
carrying capacities for same settlement. It can also
be seen that iron rods increases the ultimate load
capacities especially for 50S+50T (more than 50%)and more than 100 % for 100S. It is also noted fromthe Table 3 that clay bed reinforced with nail only
has only marginal increase of load capacity (32 %)due to skin friction along the nails. But when thesenails were used in stone columns, the load was
initially taken by the aggregates in the column at low
settlement, and at high settlement, the columns tend
to bulge, which is confined by steel nails. This
lateral confinement of nails lead to an increase of
85% for 100T and 117 % for 100S. This increase ismuch higher than 32 % increase observed in bed
with nails only. Thus, it may confirm that the steel
nails interact efficiently in stone column when it
tend to bulge laterally. It is also seen from Table 3that the capacity of steel rod-encased 100T column
is 1208 N, which is higher than the theoretical
capacity of ordinary stone column made of 100%stone aggregates (1137 N). This indicates that withsteel nail-reinforcement, complete replacement ofstone aggregates is possible with tyre chips.
For understanding the performance of stone
columns made of different mix proportions and theeffect of nail-reinforcement, a parameter:“Efficiency” is defined. Efficiency is defined as afactor: ratio of load carrying capacity of either OSC
or ESC made of any mix proportion to the load
carrying capacity of OSC made of 100S. The
comparison plot of efficiency of the stone columns
in the present study to that reported in literature isshown in fig. 13.
2.5

CONCLUSIONS
The results obtained from the model tests in thepresent study show that 100S column exhibits
highest capacity. However, it is seen that complete
replacement of the stone aggregates with shredded
tyre chips is possible when the columns are provided
with the steel rods as encasement over the entire
depth of the column. This is due to the higher
stiffness of the composite ground. The tensile
strength of steel rods also plays an important role insignificant increase in the overall stiffness.
Corrosion of the steel nails used may be a concern,
which needs to be taken care, when this technique isadopted in practice. The durability of the columns is
a concern especially when the columns are exposed
to saline water or marine and coastal water. In suchcases, galvanized rods may replace the usual steel
nails. However, field studies are essential to examine
the durability of corrosion-treated nails andgalvanized nails for confirming its field applications.

Present Study (D = 100mm)
Shivashankar et al. 2010
Efficiency = 1

2

Effciency

seen from fig. 13 that the efficiency of each of thecolumn in the present study as well as obtained fromliterature is greater than 1.0. The efficiency for 100S
stone columns measured from present study ishigher than that was measured by Shivashankar et al.
[15]. This may due to the fact that the model tests bythese authors were conducted on smaller diametercolumns and the diameter of nails were 2 mm,therefore the stiffness of nail-reinforced column
might be lesser than that was used in the present
study. IT is also seen from fig. 13 that columns
reinforced with vertical steel nails exhibit efficiency
greater than 1 even when the stone aggregates arecompletely replaced by tyre chips. Thus this ensuresthat complete replacement of stone aggregates withshredded tyre chips is possible when the columns areprovidedwithverticalnails
asencasement/reinforcement. However, there is an
issue of corrosion of these steel nails, which require
proper anti-corrosive treatment/protection whenthese nails are used in practice and it needs to be
examined through field studies. Possible use of
shredded tyre chips as backfill in retaining walls,
vibration absorbers, drainage media in landfill
leachate collection systems and as insulators beneathroads had been explored by Humphrey and his coworkers, thus it is evident that the shredded tyrechips has its applicability in wide range of civilengineering problems.

D = 60 mm

1.5

D = 75 mm
D = 90 mm

1
0.5

100S 70S+30T 50S+50T 30S+70T 100T
Mix Proportion

Fig. 13. Comparative study of efficiency of different
stone columns.

REFERENCES[1] Murugesan,
S.,
and
K.
Rajagopal,“Geosynthetic-encasedstonecolumns:
numerical
evaluation”,
Geotextiles
andGeomembranes, Volume 24, Issue 6, December
2006, pp. 349-358.

The efficiency for conventional stone columns(100S) of different diameter reinforced with steel
nails upto 3d (d = diameter of the stone column)
conducted by Shivashankar et al. [15] is also
calculated and compared with the present study. It is297

GEOMATE- Bangkok, Nov. 14-16, 2016[2] Greenwood, D. A, “Mechanical improvement ofsoils below ground surface”, Institution of Civil
Engineers Proceedings, London/UK/, 1970, pp1900.
[3] Hughes, J. M. and Withers, N. J, “Reinforcingof soft cohesive soils with stone columns”,
Ground Engineering, 1974, pp. 42-49.
[4] Bergado, D. T., Anderson, L. R., Miura, N. andBalasubramaniam, A. S,. “Soft ground
improvementinlowlandandotherenvironments”, ASCE, 1996.
[5] Alamgir, M., Miura, N., Poorooshasb, H.B. andMadhav, M.R, “Deformation analysis of soft
ground reinforced by columnar inclusions”,
Computers and Geotechnics, Volume 18, Issue
4, 1996, pp. 267–290.
[6] IS: 15284 – 2003, “Design and construction forground improvement—Guidelines. Part 1: Stone
columns”, Bureau of Indian Standards, New
Delhi, India.
[7] McKelvey, D., V. Sivakumar, A. Bell, and J.
Graham, “Civil Engineers GeotechnicalEngineering”, Volume 157, Issue GE3, July
2004.
[8] Ambily, A.P. and Gandhi, S.R, “Behavior ofstone columns based on experimental and FEM
analysis”, Journal of Geotechnical andGeoenvironmental Engineering, ASCE, Volume
133, Issue 4, 2007, pp. 405–415.
[9] Black, J.A., Sivakumar, V., Madhav, M. R. andHamill, G. A, “Reinforced Stone Columns in
Weak Deposits: Laboratory Model Study”,
Journal of Geotechnical and Geoenvironmental
Engineering, ASCE, Volme 133, Issue 9, 207,
pp. 1154-1161.
[10] Deb, K., Chandra, S. and Basudhar, P. K,“Responseof
multilayer
geosyntheticreinforced bed resting on soft soil with stone
columns”, Computers and Geotechnics, Volume
35, 2007, pp. 323-330.
[11] Malarvizhi, K. and Ilamparuthi, S. N, 2007.
“Comparative study on the behavior of encasedstone column and conventional stone column”,
Soils and Foundation, Japanese GeotechnicalSociety, Volume 47, Issue 5, 2007, pp. 873-885.
[12] Borges, J. L., Domingues, T. S. and Cardoso, A.
S, “Embankments on Soft Soil Reinforced with
Stone Columns: Numerical Analysis andProposal of a New Design Method”,

Geotechnical and Geological Engineering,
Volume 27, 2009, pp. 667–679.
[13] Gniel, J. and Bouazza, A, “Improvement of soft
soils using geogrid encased stone columns”,
Geotextiles and Geomembranes, Volume 27,
2009, pp. 167-175.
[14] Murugesan, S. and Rajagopal, K, “Studies on
the behavior of single and group of geosynthetic
encasedstone
columns”,
Journalof
GeotechnicalandGeoenvironmental
Engineering, Volume 136, Issue 1, 2010, pp.
129-139.
[15] Shivashankar, R., Babu, M. D., Nayak, S. andManjunath, R, “Stone columns with verticalcircumferential nails: Laboratory model study”,
Geotechnical and Geological Engineering,
Volume 28, Issue 5, 2010, pp. 695-706.
[16] Shahu, J. T. and Reddy, Y. R, “Clayey Soil
Reinforced with Stone Column Group: Model
Tests and Analyses”, Journal of Geotechnicaland Geoenvironmental Engineering, ASCE,
Volume 137, Issue 12, 2011, 1265-1274.
[17] Hanna, A.M., Etezad, M. and Ayadat, T, “Modeof Failure of a Group of Stone Columns in SoftSoil”, International Journal of Geomechanics,
ASCE, Volume 13, Issue 1, 2013, pp. 87-96.
[18] Serridge, C. J, “Achieving sustainability invibro stone column techniques”, Proceedings ofthe Institution of Civil Engineers-Engineering
Sustainability, Volume 158, Issue 4, Thomas
Telford Ltd, 2005, pp. 211-222.
[19] Humphrey, Dana N., Thomas C. Sandford,
Michelle M. Cribbs, and William P. Manion,“Shear strength and compressibility of tire chips
for use as retaining wall backfill”,
Transportation Research Record, Volume 1422,
1993.
[20] Ayothiraman, R., Soumya, S, “Model tests on
the use of tyre chips as aggregate in stone
columns”, Ground Improvement, Proceedingsof ICE, UK, DOI: 10.1680/grim.13.00006, 2014[21] Ayothiraman, R., Rolaniya, A. K. and
Mazumder, T, “Load-settlement behaviour ofencased stone columns made of waste tyrechips”, proceedings of 6th International
GeotechnicalSymposiumon
Disaster
Mitigation in Special Geoenvironmental
Conditions, January 21-23, 2015, IIT Madras,
Chennai, India, pp. 241-244.

298

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INCREASE OF EFFICIENCY OF CONSOLIDATION OF WEAK
WATER-SATURATED SOIL VERTICAL SANDY DRAINS AND
DINAMIC LOADINGS
Usmanov R. A., Dr. Sci. Tech., prof., vice-president of Geotechnical society of Tajikistan,
Dushanbe, Republic of Tajikistan
ABSTRACT
Results of field pilot studies on consolidation of weak water-saturated loessial soil of Central Asian regions byvertical sandy drains are given in article. In the first part efficiency of application of this method is proved in soil
conditions of the Republic of Tajikistan. In the second part features of carrying out experimental works in field
conditions, a technique of the appendix of static loadings and imitation of seismic influences are considered. In
the third part results of researches, changes of strength and deformation characteristics of the condensed soil are
analysed. In the fourth part the corresponding conclusions by results of researches are drawn. At the end of article
the bibliography is provided.
Keyword: Weak water-saturated loessial soil, Vertical sandy drains, Loading embankment weight, Static
influences, Seismic and seismic-explosive influences, Consolidation.

conditions of the Republic of Tajikistan field (natural)pilot studies which purpose was research of featuresof their consolidation by sandy drains at static anddynamic (seismic) influences have been conducted.

INTRODUCTION
Experience of construction shows that at aconsiderable thickness of a weak water-saturated
layer of earth (h > 10 m), fluidity indicator IL ≥ 0.8
and the module of deformation of soil E ≤ 5.0 MPa
the method of consolidation of soil by vertical sandy
drains from the subsequent loading condensed
thickness the weight of an embankment or othermaterials from which payload is equal is widely used
or exceeds loading from the projected construction.
In seismically active regions of the Republic of
Tajikistan, as well as in some other the Central Asian
countries, in connection with flooding of territories,
platforms of construction can be often presented by
big thicknesses of weak water-saturated loessial soil
(h = 12-20 m), with very low values of physics andmechanicals characteristics (IL ≥ 0.8; е ≥ 0.8; R ≤ 100
KPa; Е ≤ 5.0 MPa). Design and ensuring operationalreliability of buildings in the specified conditions isconnected with application of various methods ofpreparation of the artificial bases (consolidation,
fixing, replacement of weak soil, etc.) and the deviceof the deep foundations, including pile foundations.
The existing experience shows that the method of
consolidation of this soil vertical sandy drains and theloading territory [1]-[5] can be one of effective, andsometimes the only thing.
It should be noted what experience of design andmaterials of researches on consolidation of weakwater-saturated loessial soil by this method in theconditions of the republic and other countries us isn'testablished. For studying of an opportunity andefficiency of application of this method in the

TECHNIQUE OF TESTS
Researches have been conducted on an
experimental site, the weak water-saturated soil put
by big thickness with the following physicяs andmechanicals properties: density of dry soil - ρd = 1.54
t/m3, natural humidity - w = 32%, degree of humidity- Sr = 0.98, coefficient of porosity - e = 0.79,
consistency - IL = 0.8, module of deformation - E = 22.7 MPa. Level of underground waters is at a depth of
1.5 m from a day surface, and on depth more than 16
m soil is presented by water-saturated loessial loams.
Prior to the beginning of the device of sandy
drains static tests by means of round rigid stamps bysquare Ast = 1.0 m2 (dst = 1.13 m) which have been
established directly on a layer of weak soil have been
carried out (on a mark of level of underground waters).
By results of tests value of settlement resistance ofsoil of the basis has made R = 90 KPa, and thedeformation module - E = 2-2.7 MPa.
On an experimental site the ditch of 10 x 30 m insize to the level of underground waters on whichsurface the sandy pillow with h thickness h = 0.5 m
from the coarse-grained washed sand has beenarranged has been dug. This pillow provided
traficability of mechanisms on a ditch surface, and299

GEOMATE- Bangkok, Nov. 14-16, 2016

further served as a horizontal drainage for removal of
the wrung-out water from thickness of soil. The
prepared platform has been divided on 3 separate
speed up the sizes of 10 x 10 m everyone. The plan
and a section of the experimental platform is given in
Figure 1.
For comparison of results of experimental
researches the first site was loading embankment
weight without device of sandy drains in the basis, inthe second sandy drains on a grid 2 x 2 m, in the third– on a grid 3 x 3 m have been arranged. Vertical sandy
drains have been arranged with diameter of d = 0.4 m
on depth h = 6.0 m from a ditch bottom. The schemeof a vertical sandy drain is shown in Figure 2a.

made by embankment weight by dumping of layersof earth 1.0 m thick everyone, with endurance of eachstep before the conditional stabilization of
deformations accepted equal 1.0 mm/days. After aloading of thickness of soil static influences, havealso conducted researches on studying of influence ofthe seismic forces imitated by seismiс-explosive
influences.
Imitation of seismic influence has been carried out
with the help of camouflage explosions of charges of
explosive weighing 8-10 kg in (a way of detonationof electric). Charges of explosive with a gross weightof 384 kg was established in special metal containers
and plunged into boreholes on depth of 8-10 m fromthe Earth's surface. Explosive wells in number of 40
pieces have been located 3 concentric rows
(13+13+14 wells) at distance of 20, 25 and 30 m fromexperimental sites. This method is widely used inpractice of researches of the Republic of Tajikistan[6].
After completion of all works changes of physics
and mechanicals properties of soil of the condensed
thickness in laboratory and field conditions have beenstudied, and also static tests by means of metal roundrigid stamps by a standard technique are carried out.

Fig. 1 Plan and section of the experimental platform

Results of researches

For measurement of the general and layer-bylayer deformations of thickness of weak soil on depth,on all three sites special designs of superficial anddeep brands have been established (the scheme ofplacement of brands is given in Fig. 1, and theirdesigns in Fig. 2b and 2c). Superficial brands havebeen placed on a grid 3 x 3 m, and deep brands are
established through each 1.5 m up to the depth of 6.0
m. Supervision over movements of all brands werecarried out by means of leveling concerning two
motionless reference points located at distance of 40
m from the experimental platform.

Before a loading of the territory provisions of all
superficial and deep brands of rather motionlessreference points have been recorded. Loadjng of theterritory static pressure has been made by layer-bylayer dumping of a soil embankment from mix ofgravel and pebble and loamy soil by density ρ = 1.8
t/m3. When dumping an embankment there were great
difficulties in this connection only 4 layers and thegeneral static load of the condensed thickness of soil
have been poured out has made р = 75 KPa.
Dependence of deformations of the condensed
thickness of "s" on time of "t" and weight of a load of"p" for all experimental sites are given in Figure 3.

Fig. 2 Scheme of a sandy drain (a), design ofsuperficial (b) and deep brands (c): 1- prigruzochny
embankment; 2- sandy pillow; 3- sandy drain; 4weak soil; 5- reperny pipe; 6- epiploons; 7- upsetting
pipe; 8- pipe d = 76 mm; 9- metal plate

Fig. 3 Development of deformations of experimentalsites in time

Loading of experimental sites static pressure was300GEOMATE- Bangkok, Nov. 14-16, 2016

The greatest values of deformations have beenrecorded on a site where sandy drains are located ona grid 2 x 2 m (site 2). Rather good effect ofconsolidation is also gained at the device of drains ona grid 3 x 3 m (site 3). By researches it has been
established that the main part of deformations
happens in the first 5-6 days after the appendix ofeach step of loading. Apparently from the schedule
application of vertical sandy drains allows to increase
considerably quality of consolidation of thickness of
weak water-saturated loessial soil (by 5-8 times) that
testifies to rather high efficiency of application of thismethod in the conditions of weak water-saturated
loessial soil.
Features of development of layer-by-layer andrelative deformations of soil in depth of the basis of asite 2 are given in Figure 4. The analysis of a diagramof layer-by-layer movements demonstrates that themain part of deformations of soil thickness happens
at a depth up to -3.0 m where more than 70% of thegeneral draft of soil thickness are recorded. At thesame time, more than 50% of their size are the shareof depth of -1.5 m, and the deep brand located on amark of -6.0 m practically didn't move. Obviously, it
is connected with the insufficient size of the operatingpressure, as well as with the fact that on depth of thebasis the structural durability of compression of soil
increases.

circumstance is connected with increase of excessivehydrodynamic pressure in steam water which brings
to additional their wring into a body of sandy drains
and to consolidation of thickness of soil.
The following is established by the conductedpilot studies:- seismic (dynamic) influences allow to accelerate
process of filtrational consolidation and promote
improvement of quality of consolidation of thicknessof weak water-saturated soil when using verticalsandy drains;- at design of the specified method in theconditions of weak water-saturated soil and highseismic activity of territories of construction, it isnecessary to consider a possibility of additional
consolidation of thickness of soil and, therefore
developments of additional deformations infoundation of buildings when passing a seismic(dynamic) wave. Not accounting of this factor cansignificantly affect operational reliability of buildings
after their delivery in operation.
It should be noted that imitation of seismic(dynamic) influences by means of seismic-explosions
is labor-consuming and expensive action. For
acceleration of process of consolidation andimprovement of quality of consolidation of weakwater-saturated loessial soil by more expedient use ofenergy of small charges (micro-explosions) [8] whichmass for preservation of integrity of sandy drains byproduction of explosions has to be accepted within
0,3-0,5 kg is. At the same time charges of explosive
should be placed in space between drains and at 2-3
levels on height of a sandy drain (schemes of anarrangement of charges on the example of anexperimental site 2 are given in Figure 5a and 5b).

Fig. 4 Development of layer-by-layer and relative
deformations in basis depth
After completion of tests for action of static
loadings, researches of influence on the massif of thecondensed thickness of soil of the seismic loadings
imitated by seismic-explosive influences by theabove-stated technique have been conducted. At thesame time intensity of seismic influence has made 8
points (on the 12th mark scale of MSK-64) [7].
Results of tests are given in fig. 1 and show thatseismic-explosive influences promote developmentof considerable additional deformations (Sdin) which
can make up to 30% of the general deformations
recorded at a static loadings. Obviously, thisFig. 5 Schemes of an arrangement of charges of
explosive in respect of (a) and on height (b) between
sandy drains
After completion of all researches on
consolidation of loessial water-saturated soil sandy
drains, on sites 2 and 3, for definition of strength anddeformation characteristics of the condensed soil,
have carried out static tests by means of metal rigid

301

GEOMATE- Bangkok, Nov. 14-16, 2016

round stamps by square Ast = 1.0 m2 and diameter of
dst = 1.13 m. Tests were carried out by a simultaneous
loadings of two stamps located at distance not less
5,0dsh from each other through a metal platform bymeans of reinforced concrete calibrated blocks stepsof loadings on р = 25-50 KPa. In Figure 6
dependences of deformation of stamps of "s" onloading "р" are given in the condensed sites. For
comparison of the received results on the sameschedule results the stamps of tests of soil before theirconsolidation are given by sandy drains (in the natural
basis)

saturated loessial soil when using vertical sandy
drains. For their imitation by more expedient use of
micro-explosions by placement of charges of
explosive in space between drains and at various
levels on height of a sandy drain is. Besides, use of
micro-explosions will allow to increase quality ofconsolidation at rather small values of static pressure,i.e. will promote decrease in height of the poured-out
embankment (loading).
4. Results of researches also testify to need of theaccounting of seismic and dynamic influences whenusing vertical sandy drains for consolidation of weakwater-saturated soil since at the same timedevelopmentofadditional
deformations,considerable in size, in foundation of buildings after
their delivery in operation is quite possible that cannegatively affect operational reliability of buildings.
Bibliography[1] Abelev MYu, “Construction of industrial andcivil buildings on weak water-saturated soil”,
Moscow: Stroyizdat, 1983.
[2] Mangushev RA, Usmanov RA and etc.,“Methods of preparation and device of theartificial bases”, Manual, Moscow – St.
Petersburg: Publishing house of Association ofConstruction Higher education institutions, 2012.
[3] “Recommendations about rational methods ofconsolidation of weak water-saturated soil bymeans of sandy vertical drains and limy columns
for the device of the bases and the foundations,
Espoo, Finland, 1984.
[4] “Recommendations about consolidation of weaksoil temporary loading with application of sandy
and paper drains”, Yaroslavl, 1978.
[5] “Recommendations about consolidation of weakwater-saturated soil temporary loading withapplication of tape drains”, Moscow, 1985.
[6] Lekarkin VK, “A technique of research of
influence of seismic-explosive impacts on the
bearing ability of the bases”, in Proc. of theinternational scientific conference "Modern
Aspects of Development of Aseismic
Construction and Seismology", Dushanbe, 2005.
[7] Set of rules of the joint venture 14.13330.2014"Construction in seismic countries", thestaticized edition Construction Norms andRegulations of II-7-81*, Moscow, 2013.
[8] Ivanov P L, “Consolidation of low-coherent soil
explosions”, Moscow: Nedra publishing house,
1983.

Fig. 6 Dependence a deposit of stamps from loadingon the condensed bases
The received results demonstrates that
consolidation of thickness of weak soil with use ofvertical sandy drains allows:- to increase values of settlement resistance (toincrease the bearing ability) basis soil to R = 250 kPa
that almost by 3 times exceeds these values before
consolidation;- to increase values of the module of deformationof the condensed soil to 4 times;- to increase values of specific coupling of thecondensed soil twice (by results of static sounding).
Conclusions
1.
In general, the conducted pilot studies testify toefficiency of application of vertical sandy drains forconsolidation of the bases put by water-saturated
loessial soil, even at small sizes the static loadings.
2.
The main lack of this method is complexity of
dumping of a prigruzhayushchy embankment of bigheight in the extensive territory that demands use of
considerable volume of soil and labor expenses.
Therefore it is necessary to find more effective waysof a prigruzka of thickness of soil (for example, to use
a zagruzheniye of the condensed territory water
weight by analogy with hydrotests of metal tanks forstorage of oil and oil products).
3. Seismic (dynamic) influences significantly
increase efficiency of consolidation of weak water-302

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INFLUENCE OF CLAYSTONE DETERIORATION ON SHEAR
STRENGTH OF BACKFILL
1

Pisut Rodvinij1 and Pitiwat Wattanachai1
Department of Civil Engineering, Chiang Mai University, Thailand
ABSTRACT
Claystone in Mae Moh mine after excavation has been used as a backfill material to counterweight instability
slope. However, claystone deteriorates when it has been exposed to wet-dry process in the natural condition. The
investigation on the properties of claystone under natural condition and control condition can clarify thedeterioration effect due to wet-dry process. Deterioration can be defined as the alteration process on size
distribution and strength of claystone. Sieve analysis test and unconfined compression test were conducted on
claystone samples to investigate the deterioration effect on size distribution and strength of claystone. Claystone
exposed to natural condition having a decrease of an average particle size (D50) to be approximately 50 mm in 12
weeks and unconfined compressive strength of claystone was decreased 45% in 3 weeks. In contrast, claystoneunder control condition gave the same size distribution and unconfined compressive strength slightly decreased.
Thus, the result demonstrated wet-dry process in natural condition was the main cause of claystone deterioration.
The decrease on size distribution of claystone due to deterioration can be the cause of reduced on backfill shear
strength. In order to clarify this assumption, direct shear test was conducted on 3 different particle sizes ofclaystone to simulate the effect of size reduction on shear strength of backfill. The experimental results
demonstrated shear strength of the larger particle size samples was higher than the smaller especially under highnormal stress. In conclusion, the deterioration of claystone significantly affects on shear strength of backfill.
Keywords: claystone, deterioration, shear strength, size distribution, unconfined compressive strength

the alteration process of claystone occurring under
the direct influence of atmosphere and hydrosphere[3]. The alteration process can be in form of either
physical and/or chemical deterioration. The physicaldeterioration causes disintegration (or breakdown) oforiginal fabrics and also imposes new fabric features.
While the chemical deterioration, which usuallytakes a long time, forms discoloration of the affectedrock and change in mineralogy [4].

INTRODUCTION
Study area located at the northeast side of the pit
wall in Mae Moh mine, called area 4.1. This area
will become problematic because a normal fault
separated the layers have been displaced in a dip
direction to the normal fault. Fig. 1 shows a crosssection of area 4.1. According to the mine plan, K
and Q seams (green layer and red layer in Fig.1) oflignite must be excavated [1]. K and Q lignite layers
have supported the potential mass (gray layer in
Fig.1). Thus, after excavating K and Q seams oflignite, the potential mass over the lignite layer
becomes a sliding. The suitable method to solve thisproblem is cut and fill method.
The cut and fill method was used to excavatelignite in this area. The area will be mined andpartially undercut. Thus, the G1 weak planeinterface is also partially daylight. After completionof mining, backfilling will begin to support the slope
face and maintaining adequate room for mining
activities. The mining- backfilling cycle is iterative
until all the lignite in area 4.1 is completely minedout. It is necessary to fill the current pit and cut theneighboring slope in subsequent procedure [1], [2].
Claystone has been used as a backfill material tocounterweight and supported the potential mass.
However, claystone deteriorated when it has been
exposed to weather change. The deterioration ofclaystone can be the cause of reduced in the stabilityof backfill slopes. Deterioration can be defined as

Fig. 1

Cross section plot of area 4.1 [1]

In its fresh condition, claystone usually stiff orhard and quite difficult to excavate without
mechanical equipment. The occurrence of physicaland chemical changes in deterioration process canlead to significant reduction in strength value [5].
Moreover, claystones have specific characteristics,
mainly found to exhibit slake-deterioration within a303

GEOMATE-Bangkok, Nov. 14-16, 2016short period of time when exposed to the atmosphere
and/or moistened. This will induce some significant
problems in varied engineering activities [6].
Shear strength behavior of claystone backfill was
similar to rockfill or granular soil that most dependon friction angle [7]. Thus, the decrease on size
distribution of claystone due to deterioration can belead to the reduced friction angle of backfill.
Eventually, stability of claystone backfill can be
reduced due to deterioration of claystone.

disintegrated into small pieces. When the samples
were in natural condition during 6-9 weeks, almost
all of claystone samples was broken down into verysmall pieces and could not classify the pieces of thesample before breakdown.

Initial1 weeks

3 weeks

6 weeksSCOPE
The investigation on claystone properties under
the natural condition and control condition cancharacterize the deterioration behaviors of claystone.
Natural condition, claystone samples were exposedto the natural weathering process. In contrast,
claystone samples under control condition werelocated in laboratory room (The temperature was
between 25-28 C) and they were not exposed to theweather change. The alteration process of claystoneunder two different conditions can be suggested thatwet-dry process in natural condition was the causesof claystone deterioration. The alteration process on
the size distribution of claystone was investigated bysieve analysis test. The alteration process on thestrength of claystone was investigated by unconfinedcompression test.
The influence of claystone deterioration to theshear strength of backfill was investigated by shear
strength test. Direct shear test was conducted on 3
different particle sizes of claystone samples in order
to simulate the decrease of claystone particle sizedue to the deterioration affected shear strength of
backfill.

9 weeks

Fig. 2

Deterioration effect on characteristic ofclaystone under natural condition

THE EFFECT OF DETERIORATION ON SIZE
DISTRIBUTION OF CLAYSTONE
The physical characteristic of deteriorated
claystone demonstrated breakdown of claystonewhen exposed to natural weathering processes. Size
distribution of claystone backfill was changed due todeterioration effect. In order to investigate thedeterioration effect on size distribution of claystone.
Sieve analysis was conducted on claystone samples
in a natural condition and a control condition during
1-12 weeks to clarify the alteration of size
distribution.
Fig. 3 indicated the size distribution of claystone
decreased under natural environment during 1-12
weeks. Size distribution of the samples wasdecreased with increasing elapse time. Considering
the average particle size (D50) of claystone samples,D50 of claystone samples decreased from 60.0 mm
remaining to 9.0 mm in 12 weeks. The deterioration
greatly affected during 2 weeks as D50 of thesamples in 2 weeks was decreased from the firstweek approximately 40 mm. After 2 weeks, theaverage particle size of claystone was slightly
decreased as D50 of the samples was decreased
approximately 11 mm during 2-12 weeks.
In contrast, claystone under control condition
gave the same size distribution when the time passed
as shown in Fig 4. The results can be concluded thatclaystone deteriorated, specifically under natural
condition. Thus, wet-dry process in natural condition
was the main cause of claystone deterioration.

VISUAL CHARACTERISTIC OF
DETERIORATED CLAYSTONE
Physical characteristics of the deteriorated
claystone can be observed by visible. Claystone
samples were exposed to the natural weathering
process during 0-9 weeks. Claystone samples
dimension between 10-25 cm were collected fromMae Moh mine. The deterioration effect to acharacteristic of claystone in natural condition wasshown in Fig. 2 When time passed, the deterioration
effect was demonstrated by breakdown of original
claystone.
Physical characteristics of original claystone are
generally gray and it is relatively stiff or hard. Afterclaystone exposed to natural weathering process in 1
week, it was starting to intensively slake andgenerally followed by noticeable heaving. Some ofclaystone samples were started break to small
particles and showed some visible fractures on thesamples. In 3 weeks, the visible fractures onclaystone were increased and most of the samples
304

GEOMATE-Bangkok, Nov. 14-16, 2016

Fig. 5
Fig. 3On the other hands, the strength of claystone
samples under control condition decreased from 2.8
MPa remaining to 2.4 MPa or reduced
approximately 15% in 4 weeks. After 4 weeks, thestrength of samples was slightly decreased.
Unconfined compressive strength was remaining
2.1 MPa in 12 weeks. Hence, unconfinedcompressive strength of claystone samples in control
condition has decreased approximately 25% during
12 weeks.

Size distribution of claystone samples undernatural condition

Fig. 4

Unconfined compression test on claystone

Size distribution of claystone samples undercontrol condition

THE EFFECT OF DETERIORATION ON
STRENGTH OF CLAYSTONEIn fresh condition, claystone is usually stiff orhard materials. However, the deterioration process
can lead to significant reduction in strength value [5].
In order to investigate the deterioration effect onstrength of claystone, unconfined compression test
was conducted on claystone samples under a natural
condition and a control condition during 1-12 weeks
to reconfirm a strength reduction of claystone. Sizeof trimmed claystone specimens that used in this test
was 5 cm in diameter and 10 cm height as shown in
Fig 5.

Fig 6.

The results of unconfined compressive test

Regarding the age of samples at 3 weeks,
unconfined compressive strength of the samples
under natural condition reduced approximately
45% but unconfined compressive strength of thesamples
undercontrolcondition
reducedapproximately 10%. Thus, the strength reduction
rate of claystone under natural condition was higher
than claystone under control condition. The results
can demonstrate the influence of wet-dry process
due to weather change extremely affected thestrength of claystone.

The experimental results of the deterioration
effect on a strength of claystone was shown in Fig.

SHEAR STRENGTH OF CLAYSTONEBACKFILL

6. Unconfined compressive strength of claystone
samples under natural condition decreased from 3.2
MPa remaining to 1.8 MPa or reduced
approximately 45% in 3 weeks. After 3 weeks could
not be prepared the specimens for the dimension thatused in the test because claystone specimens
deteriorated into small pieces.

Shear strength of claystone backfill was mostly
depended on friction angle of materials. Then, thedecrease of size distribution due to the claystone
deterioration was lead to reduce shear strength of
backfill. In order to verify the decrease of particle
size affect on shear strength of backfill, direct shear
test was conducted on the various particle sizes of305

GEOMATE-Bangkok, Nov. 14-16, 2016

claystone samples. The experimental results express
the deterioration effect on shear strength of backfill.

line in the high normal stress (254-930 kPa). The
higher effective normal stress will yield lower valuesof friction angle.
Fig 9 shown the reduction of particle size
affected shear strength of backfill. Shear strength
envelope of the samples at D50 equal to 19.0 mm
was higher than the samples at D50 equal to 9.5 mm
and 2.0 mm, especially under high normal stress.
Under low normal stress, shear strength envelope ofthe samples at D50 equal to 19.0 mm, 9.5 mm and
2.0 mm was similar. The maximum shear stress ofthe samples at D50 equal to 19.0 mm was higher than
the samples at D50 equal to 9.5 mm and 2.0 mm
approximately 120 kPa and 140 kPa respectively.

Table 1 Size of claystone samples
D50
(mm)

Size of samples
Passing sieve 25.4 mm retained on sieve
19.0 mm (50%) + Passing sieve 19.0 mm
retained on sieve 12.7 mm (50%)
Passing sieve 12.7 mm retained on sieve 9.5
mm (50%) + Passing sieve 9.5 mm retainedon sieve 4.75 mm (50%)
Passing sieve 4.75 mm retained on sieve
2.0 mm (50%) + Passing sieve 2.0 mm
retained on sieve 0.85 mm (50%)

19.0

9.5

2.0

Direct shear test was carried out on the samples
to obtain shear strength properties under
unconsolidated undrained condition. Size ofclaystone specimens was prepared according to table
1 and normal stress that used in this test was 0-930
kPa. Direct shear test was conducted on 6 inches indiameter of specimen by rock direct shear testingmachine as shown in Fig 7, Fig 8 and compacted
specimens at moisture contents same as sampled in
the field (moisture contents 26%).
Fig. 9Size of claystone affect to shear strength of
backfill

The Influence of deterioration affects to shear
strength parameter
D50=19 mm

Fig. 7

D50=9.5 mm

D50=2.0 mm

Shear strength parameter of claystone backfill
samples at various particle sizes which were
cohesion and friction angle was shown in Table 2.

Claystone samples prepared for direct shear
test

Table 2 Shear strength parameter of the samples
D50
(mm)
19.0
9.5
2.0

Fig. 8Low NormalStress (0-254 kPa)c (kPa)ϕ( ̊ )
0
44.1
0
43.8
0
43.1

High Normal Stress(254-930 kPa)c (kPa)ϕ( ̊ )
53
35
103
27
109
26

Table 2 shown friction angle of the samples at
D50 equal to 19.0 mm was more than D50 equal to
2.0 mm approximately 1° under low normal stress.
Under high normal stress, friction angle of thesamples at D50 equal to 19.0 mm was more than thesamples at D50 equal to 9.5 mm and 2.0 mm
approximately 8° and 9° respectively. Cohesion of
the samples at D50 equal to 19.0 mm was less than
the samples at D50 equal to 9.5 mm and 2.0 mm
approximately 50 kPa and 56 kPa respectively.

Rock direct shear testing machine

The reduction of claystone particle size affects on
shear strength of backfill
Shear strength characteristic of claystone backfill
was similar to granular soil or rockfill. Thus, shear
strength behavior is actually nonlinear. Thisnonlinear can be approximated a straight line in therange of low normal stress (0-254 kPa) and a straight306

GEOMATE-Bangkok, Nov. 14-16, 2016

strength of backfill reduced by the decline ofclaystone particle size, especially under high normal
stress. Overview of this study can be concluded thatclaystone deteriorated under natural condition due towet-dry process and the deterioration of claystone
significantly affects to decrease shear strength of
backfill.
REFERENCES[1] Leelasuksaree C, Mavong N, Khosravi M H,
Pipatpongsa T, "Physical and Numerical
Models of Undercut Slope Lying on Steeply
Inclined Bedding Plane", 17th National
Convention on Civil Engineering, 2012, pp.
GTE045-1 -GTE045-10.
[2] Mavong N, Tunkeaw S, Mungpayabal N,
Boonsritun N, "Mining Plan and Progressive
Excavation in Lowwall Area 4.1 of Mae Moh
Mine", International Symposium on Earthquake
Hazard Potential and Preparedness for Safety in

Fig. 10 Relationship between friction angle and D50
Fig. 10 shown the correlation between friction
angle and D50 of the samples. Friction angle reducedby the decline of claystone particle sizes. In therange of low normal stress, friction angle wasslightly decreased when the particle size of claystone
was smaller. In contrast, friction angle in the rangeof high normal stress was rapidly decreasing when
D50 of the samples equal to 19.5 mm reduced to 9.5
mm. However, the average particle size of thesamples which lower than 9.5 mm, a trend of friction
angle was slightly decreased. The results can be
described the decrease of claystone particle size due
to deterioration was a cause of reduced backfill
shear strength especially under high normal stress.

Coal Mining, 2011.

[3] Sounders M K, and Fookes P G, "A review of
the relationship of rock weathering and climate
and its significance to foundation engineering",
Eng. Geol, 4, 1970, pp. 289-325.
[4] Sadisun I A, Shimada H, Matsui K,"Characterization of Weathered Claystone andTheir Engineering Significance", Indonesian
Scientific Meeting 2000, 2000, pp. 1-8.
[5] Sadisun I A, Shimada H, Matsui K,"Determination of Strength of Subang
Formation Claystone Due to Weathering",
Proceeding of the 3rd Asian Symposium on
Geology and the Environment (ASEGE), 2001,
pp. 37-46.
[6] Sadisun I A, Shimada H, Ichinose M, Matsui K,"Study on the physical disintegration
characteristics of Subang claystone subjected tomodified slaking index test", Geotechnical andGeological Engineers, 2005, pp. 199-218.
[7] Das Braja M, "Shear Strength of Soil",
Advanced Soil Mechanics, Singapore :McGraw-Hill, 1985, pp. 402-416.
[8] Budhu M, "Some Causes of Slope Failure", Soil
Mechanics and Foundation, New Jersey : John
Wiley & Sons, 2000, pp. 525-528.
[9] Mitchell J K, Soga K, "Physical Process of
Weathering", Fundamentals of Soil Behavior,
New Jersey : John Wiley & Sons, 2005, pp. 8.
[10] Naderian A R, William D J, "Bearing Capacityof Open-Cut Coal-Mine Backfill Material"
Transactions of the Institution of Mining andMetallurgy, Section A-Mining Industry, 1997,
pp. A30-A34.
[11] Shakoor A, Gautam T P, "Assessing the Slaking
Behavior of Clay-Bearing Rock", 10th Annual
Technical Forum Geohazards Impacting
Transportation in the Appalachian Region, 2010.

CONCLUSION
Claystone has been used as backfill material tocounterweight and supported instability slope in
Mae Moh mine. In its fresh condition, claystone
usually stiff or hard. However, claystone
deteriorated when it has been exposed to weatherchange. The deterioration of claystone can be thecause of reduced in the shear strength of backfill.
Deterioration can be defined as the alteration
process on size distribution and strength ofclaystone. The investigation on claystone properties
under natural condition and control condition cancharacterize the deterioration behaviors of claystone.
Size distribution and strength of claystone undernatural condition decreased due to deterioration
effect. In contrast, claystone under control condition
can maintain size distribution and strength. Thus, theinvestigation on claystone properties suggested thatthe deterioration effect occurs only under natural
condition due to claystone has been exposed to wetdry process.
Stability of claystone backfill slope depended on
shear strength of backfill material. The reduction on
size distribution of claystone due to deterioration canbe the cause of reduced backfill shear strength.
Therefore, shear strength test was conducted on 3
different particle sizes of claystone to simulate theeffect of size reduction on shear strength of backfill.
The experimental results demonstrated shear307

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

FINITE DIFFERENCE ANALYSIS OF A CASE STUDY OF VACUUM
PRELOADING IN SOUTHERN VIETNAM
1,2

Cong-Oanh Nguyen1, Thanh Thi Tran2 and Van-Tram Thi Dao3
Division of Soil Mech. & Foundation Eng., Southern Institute of Water Resource Research,
Vietnam; 3Faculty of Civil Engineering, Transportation College 3, Vietnam
ABSTRACTThe vacuum preloading method was first developed in 1952, since then it has become a commonly usedmethod all over the world in general and in Vietnam in particular to accelerate the consolidation of the soft
ground. However, there is still misunderstanding of the variation of vacuum pressure over the depth ofprefabricated vertical drain (PVD) among researchers. The paper presents a case study of soft ground
improvement by vacuum preloading method for a land development project in Sothern Vietnam. A series offinite difference analysis (FDA) of the improved ground with multi-layer models to get insight into theeffectiveness of the method are conducted. Input parameters are characterized from constant rate of strain (CRS)
consolidation tests on undisturbed samples retrieved by stationary piston sampler. Vacuum pressure isconsidered as equivalent surface load in the analyses. In comparison with field monitored data, the analyses
show good agreement for magnitude vs. time for both settlement and dissipation of excess pore water pressure aswell as strength gain for the first 10-m soft ground while the bottom part shows little or no improvement.
Keywords: Consolidation, CRS, FDA, PVDs, settlement, soft clay, un-drained strength, vacuum preloading

INTRODUCTION
PVDs despite the length of PVDs may exceeds theeffective depth of vacuum distribution in theformulation of the analytical equations, which is notcomplied with the Pascal’s law for the cases of
PVDs deeper than 10 m [6], [7] [8], and [9].
It had been stated [10] that below z 0 , there is noeffective vacuum pressure, which confirm the workof Pascal’s law on the distribution of vacuum
pressure over the soft soil deposit and PVDs length.
This means that there is still disputed research on the
distribution of vacuum pressure over the improved
zone. However, this issue has not been clearly
confirmed by full scale testing of the method. The
paper presents a case study of soft ground
improvement by vacuum preloading assisted bysurcharge and PVDs in the Southern area ofVietnam by simulation of the one dimension
problems with multi-layer soil model in a finite
difference code named CONSOPRO [11]. A newconcept of vacuum pressure distribution over thesoft soil deposit and PVDs is proposed. Various
finite difference analyses (FDA) is carried to provethe correctness of the proposed distribution.

Recent advances in ground improvement all over
the world allow geotechnical engineers to deal withthe soft soil treatment with large thickness in general.
Vietnam is located in the Holocene soil deposit fromthe North to the South part of the country especiallyin Red River and Mekong Delta. Among the mostpopular soft ground treatment methods, vacuum
preloading assisted with prefabricated vertical drains(PVDs) has been increasingly utilized for the recent
years.
First introduced by Kjellman [1], the method of
vacuum preloading with and without surcharge
assisted by PVDs has become one of the mostpopular way of ground improvement over the soft
clayey deposits [2]. The use of PVDs is to reduce thedrainage path length to speed up the consolidation
process which may take many years to complete [3]& [4]. With the use of surcharge the soft grounddeposit may undergo long time of construction andrisk of instability and there has been lack of sandsupply source, therefore vacuum has been applied inconjunction with surcharge to reduce the volume ofsurcharge sand and has proved the efficiency of thesoft ground improvement work [2] & [5]. Even there
has been development in theory and application of
vacuum preloading over the past few years, there is
still misunderstanding and misleading of the use of
the vacuum distribution over the whole length of

THE SITE AND SOIL PROFILE
The site is located in the South part of Vietnam
and within Mekong delta as indicated in Fig. 1. The
considered area is 100 m by 250 m. The area is308

GEOMATE- Bangkok, Nov. 14-16, 2016

state is average of 0.6 for the top 3 m of soft clay
while the below show values ranging from 0.9 to 1.1.

P-15

200

P-20
VG-13
P-16
CPTu-5

VG-12

Reclamation

VG-15
EX-06

CPTu-2P-14

50
Vietnamese
Hoang Sa Islands0

FVT-01
CPTu-01
CRS borehole (Before vac. pre.)055

Fig. 2
Vietnamese
Truong Sa Islands

Fig. 1110-35

γ(kN/m3)

100 200 10

50 100 0

0

0

15

20

Surcharge

Surcharge

Surcharge

Reclamation Fill

Reclamation Fill

Reclamation Fillσy'(CRS)

Elevation (m)-25Layout of instruments and investigation

5

The area was designed to be improved by
vacuum and surcharge preloading to reduce post
construction settlement and to ensure the slope
stability during the construction of the surcharge fillby hydraulic method. Prefabricated vertical drains(PVDs) were used in conjunction with thepreloading method the acceleration of theconsolidation of the soft clay deposit.
Even there was geotechnical investigation donein 2014 for Front-End Engineering Design (FEED),the contractor was requested to do the additional soil
investigation just before the start of the soil
improvement work to determine the compressibility
and consolidation characteristics of the soft soil
deposit in order to estimate the ground settlement
under surcharge load and during the time ofoperation. For the considered area, there was one
borehole CRS with application of the stationary
piston sampler to avoid disturbance to the soft
samples, one electrical vane shear test FVT-01 andone electrical piezocone penetration test CPTu-01 as
indicated in Fig. 2. The soft soil has natural watercontent close to liquid limit and ranging from 50 %to 75 % in Fig. 3(a). The thickness of the deposit isaround 16 m. CRS consolidation tests are used todetermine the compressibility indices andconsolidation characteristics of the soft ground withthe application of the method in [11]. The
compressibility indices are presented in Fig. 3(c).
The compressibility indices in normally consolidated

-20

-30

Site location

LL, PL & wc (%) σy' (kPa)Site location

-10
-15

P-18

P-13 CPTu-3-5Soft clay

1-5

2
σv0'+20 kPa-10Extensometer
Piezometer

The site0

Extensometers - EX

PBH-02 PZ-06 P-19
VG-11
VG-16
P-17
EX-04
PZ-04
VG-17
VG-10

100

5

Surcharge fillqT - σv0

150

VG-14

Piezocone penetration

EX-05
PZ-05
CPTu-4

Settl. Plate - P
Ext. - EX
Piez. - PZ
Borehole, CPTu, FVT (Before vac. preloading)
Borehole, CPTu, FVT (After vac. preloading)
Vac. gauge - VG
CPTu (After vac. preloading)Elevation, (m)NORTH

Piezometers - PZ

250

Distance in meters

around 30 km from the East Sea, which is affectedby fluctuated seasonal tide. The original groundelevation is lower than the average sea level.

1
2

3
3
4

-15

4
5

-20
(a)

-25

wc
LL
PL

σv0'
w'o filled sandVoid ratio e0 Cc1, Cc2 & Cr
5

Elevation (m)00cv (cm2/d)

0.6 1.2 101 102 103

1.2 2.4 0
Surcharge

Surcharge

Surcharge

Reclamation Fill

Reclamation Fill

Reclamation Fill

-5
-10
-15
-20
-25

Fig. 3

309

6
(c)

(b)

(d)

(e)

Cc1
Cc2
Cr

(f)

Soil profile and consolidation characteristics

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Compressibility and consolidation characteristics of soft soil sub-layers

Void Ratio, eSample Depth (EL.)
Ho
σ' v0
e0
σ' y
σ' b
Cr
C c1
C c2
0
N
m (mCD)
(cm)
(kPa)
(kPa)
(kPa)
1
3.9 (-1.1)
160
3
1.566
26
400 0.050
0.42
0.42
2
5.9 (-3.1)
200
16
1.565
27
175 0.050
0.65
0.60
3
7.9 (-5.1)
200
29
1.472
46
300 0.075
0.52
0.50
4
9.9 (-7.1)
200
39
2.037
53
200 0.075
0.95
0.80
5
11.9 (-9.1)
200
52
1.598
110
200 0.075
0.95
0.95
6
13.9 (-11.1)
200
63
1.816
89
200 0.075
1.10
0.90
7
15.9 (-13.1)
200
74
1.823
93
200 0.075
1.05
0.95
8
17.9 (-15.1)
200
86
1.681
110
200 0.075
1.05
0.90
9
19.9 (-17.1)
100
103
0.852
125
300 0.050
0.25
0.25
Notes: The coefficient of consolidation in OC is 10 times of that in NC state c v(OC) =10c v(NC) .
2.4

INSTRUMENTATION

2.0

The monitoring instruments consist of 8 surface
settlement plates from P13 to P20, 3 piezometergroups (PZ-04 to PZ-06) of 4 tips at each group; 3extensometer groups (EX-04 to EX-06) of 9
magnetic rings at each group and 8 vacuum gauges
from VG-10 to VG-17 as indicated on Fig. 2. The
average installation elevation of surface settlement
plate is +2.700 (mCD). The installation elevation ofpiezometer tips and extensometer rings is presented
in Table 2.

1.6
1.2
0.8
0.4
104

c v (cm2/d)c v(NC)c v(OC)2(cm /d)
7000
700
450
45
500
50
500
50
300
30
270
27
220
22
350
35
5500
550

103
102
101

Fig. 4

Table 2 Elevation of PZ tips and EX magnetic rings
101

102
σ'v (kPa)

EX-04-2.37 (mCD)-5.99 (mCD)-9.67 (mCD)-13.31 (mCD)-17.17 (mCD)-21.08 (mCD)-25.07 (mCD)-28.90 (mCD)-32.01 (mCD)PZ-04-3.38 (mCD)-7.38 (mCD)-11.38 (mCD)-15.38 (mCD)103

CRS consolidation test results

Consolidation yield stress varies as much as
linear over depth, which is shown in Fig. 3(b). Preoverburdened pressure POP, which is the difference
between consolidation yield stress σ’ y andoverburdened pressure σ’ vo is found to be 20 kPa
and OCR larger than 1.0 for the site. This confirmsthat soft clay deposit is in slightly over consolidated
state and in consistence with data found in [12] &[13], which stated that there is no Holocene deposit
in Mekong Delta having OCR less than 1.0. Voidratio for the whole clay layer is larger than 1.5,
resulted in high compressibility of the clayey soil asplotted in Fig. 3(d). Summary plots of CRS test
results are shown in Fig. 4 upon which thedetermination of compressibility and consolidation
yield stresses is complied with the method found in[11]. Determined values of the soft clay properties
are tabulated in Table 1.

EX-05-2.43 (mCD)-6.01 (mCD)-9.75 (mCD)-13.54 (mCD)-17.34 (mCD)-21.21 (mCD)-25.11 (mCD)-28.96 (mCD)
N/A
PZ-05
-3.43 (mCD)-7.43 (mCD)-11.43 (mCD)-15.43 (mCD)

EX-06-2.15 (mCD)-5.98 (mCD)-9.85 (mCD)-13.75 (mCD)-17.68 (mCD)-21.61 (mCD)-25.57 (mCD)-28.91 (mCD)-31.93 (mCD)PZ-06-3.39 (mCD)-7.39 (mCD)-11.39 (mCD)-15.39 (mCD)

The piezometer tips were all installed within thesoft clay deposit while extensometer rings were
installed down to the elevation of -30.xx (mCD), of
which 5 elevated rings are within the soft clay
deposit. Vacuum gauges were installed underneaththe impervious membrane in order to monitor the310

GEOMATE- Bangkok, Nov. 14-16, 2016

difference between underneath the membrane andoutside air. After the vacuum pressure was measured
stably within the targeted range from 70 kPa to 80
kPa as average values, the surcharge fill was loaded
in a ramp increase until the thickness reached 1.5 m.
The design dwelling time is 160 days from the startof site reclamation.

vacuum pressure transmitted from pumps to thedrainage sand fill.
Correction of the settlement of piezometer tips istaken into account to evaluate the excess pore waterpressure measurement. The piezometer tip elevation
is different from that of extensometer rings;
therefore interpolation of measured settlement ofthese rings is used with consideration of initialinstallation elevation of the tips and magnetic rings.
In order to calculate the excess pore water pressure,the ground water elevation, which is related to the
reduction in level during operation of vacuum
pumps in the treated area, is considered. The groundwater level inside the vacuum treated area can becalculated by using the initial ground water level tosubtract the reduction of its elevation caused by theeffect of vacuum pump operation.

New Procedure for Vacuum LoadCONSOLIDATION ANALYSIS
Analysis Method and Software

5

The authors apply the finite difference method
and the coded software which has been coded andcopyrighted by the authors. An axisymmetric unit
cell around the vertical drain is considered in theanalysis. As soon as the vertical drains are installedin the ground, the consolidation boundary condition
is introduced at the vertical drain. The finite
difference code named CONSOPRO version 1.0
allows users to consider up to 20 subsoil layers and
50 loading steps including ramp and sudden increasein loading, which is the same procedure applied in[11]. There is no clear data about the correlation
between the vertical and horizontal coefficient ofconsolidation; therefore, assumption of the ratio ofhorizontal values over vertical values is 3, whichmeans c h =3c v which was reported in [11].

Elevation (m)0040 800su (kPa)40 800 40 800

40 80

Ground elevation + 2.800
Surcharge fill

Surcharge fill

Surcharge fill

Extensometer
Piezometer

Fig. 5 shows the comparison between fieldtesting before and after ground improvement. Thereis considerable increase in un-drained shear strengthfrom the top of soft clay down to elevation of -8 m
Chart Datum Level (mCD) while there is almost no
strength gain for the deposit below -8 m (mCD),
which means that the effectiveness of vacuum
preloading could be within 10 m down from thevacuum pump elevation.

Surcharge fill

1
2

-5

1
2

3

-10

4

3
4

5

-15

6
CPTu-02

CPTu-03

CPTu-04

CPTu-05-20(qT - σv0)/16 (CPTu-01, before vac. pre.)FVT-01 (before vac. pre.)FVT-02 (After vac. pre.)(qT - σv0)/16 (CPTu, after vac. pre.)Torvane (after vac. pre.)Fig. 5

Piezocone after vacuum preloading

Therefore, the authors propose vacuum loading
distribution over depth as the following:
1. Constant distribution of vacuum loading frompump elevation down to the effective depth
which does not exceed 10 m. This means that
the Pascal law is true;
2. Linear distribution of vacuum preloading fromthe effective depth to the maximum depth which
follows Pascal law;
3. The effective depth is dependent on vacuum
pump efficiency;
4. The vacuum load is equivalent to the surcharge
load in the same amplitude;

Construction Sequences
From the original ground elevation of -0.8(mCD), a reclamation fill was conducted byhydraulic method. The fill thickness was averaging
of 2.8 m. A drainage sand fill of 0.5 m, which isused for acceleration of the drainage of pore water inthe ground, was carried after the reclamation fill.
Then prefabricated wick drains were installed bystatically pressing with the rectangular mandrel size
is 60 mm by 120 mm in 1 m by 1 m square grid
pattern. Impervious HDPE membrane was installed
to cover all the treated area, then vacuum pumps
were started to generate the atmospheric pressure

Consolidation Analysis
Consolidation analyses are carried out at themonitoring instruments. The proposed distribution
311

GEOMATE- Bangkok, Nov. 14-16, 2016of vacuum pressure vs. depth within the effective
zone, which accounts for Pascal law, is applied inand c h =3c v which was reported [11] is used.

the calculation. Input parameters are presented in

Settlement (cm) E.L. (m)E.L. (m)Settlement (cm)

0
0
50
100
150
200
250
4
2
0
0
50
100
150
200
250
50

PZ-06
+90 kPa

P-013

P-014

P-015

P-013

P-014

P-015

P-016

P-017

P-018

P-016

P-017

P-018

50

100

P-019

P-020

P-019

P-020

150

20050

100

150

100

150

200

200

t (days)Fig. 7 Analysis of surface settlement+80 kPa

E.L. (m)Field monitoring data

+80 kPa+15 kPa
150

Calculated by CONSOPRO

4
2
EX-06

EX-05

EX-04
0
0

STOTAL (cm)

PZ-04
PZ-05
0
80
+78 kPa+80 kPa40
0-40 -3.38m-3.43m
-3.39m-80
80+80 kPa+70 kPa40
0-40 -7.38m-7.43m
-7.39m-80
80
40
0-40 -11.38m +45 kPa-11.39m-11.43m +40 kPa
-80
80
40
0
-40 -15.38m +15 kPa-15.39m-15.43m +30 kPa-80
100
100 150 20050
50
100 150 20050
t (day)

+20cm

+20cm

+20cm

From sensor 1

From sensor 1

From sensor 1

50
100
150

200

200
0

Analysis of excess pore water pressure

S (cm)

E.L. (m)EPWP (kPa) EPWP (kPa) EPWP (kPa) EPWP (kPa)

2

Fig. 62

Calculated by CONSOPRO

4

0
0
50
100
150
200
250
4

E.L. (m)Fig. 6 shows the results of finite difference
analysis of excess pore water pressure dissipation at
3 piezometers PZ-04, PZ-05 and PZ-06. A
correction of pressure, which is equivalent to thevacuum pressure distribution over depth, is applied.
The results indicate that the distribution of vacuum
pressure from the first to the second piezometer tip
is almost constant and ranges from 70 kPa to 90 kPa;
however, below the elevation of -7.30, the vacuum
pressure reduce considerably.

2

Settlement (cm)

Excess pore water pressure

Field monitoring data

Calculated by CONSOPRO

Field monitoring data
4

Down to -6m

Down to -6m-6m to -10m

-6m to -10m

-6m to -10m

-10m to -13m

-10m to -13m

-10m to -13m

+5cm

+5cm

Below -13m
100
50
100 150

Below -13m
100 15020050
t (day)Below -13m
100 150
20050

S (cm)

50
100
0

S (cm)+15cm

Down to -6m

1000

The reduction of vacuum pressure for deeper
zone of the soft clay can reach 50% at depth deeper
than -7.30 and 80% for elevation deeper than -15.30.
It is clear that the vacuum pressure is not transmitted
constantly over the depth of prefabricated verticaldrains (PVDs) in such cases where PVDs are
installed deeper than 10 m from the vacuum pumpbase elevation. The vacuum pressure distribution
over depth from piezometer data is consistent withthose from piezocone penetration data presented inFig. 5.
The plots show that the field monitoring data ofexcess pore water pressure are consistent with thecalculated values. Right at the time (t=xx days) ofstarting vacuum pump, the excess pore waterpressure increase rapidly and followed by thedissipation process.

+15cm

+20cm
50

50

S (cm)

100
050

Fig. 8200

Analysis of multi-layer settlement

Surface and multi-level settlement
The results of finite difference analysis shows, inFig. 7, the consistence between the calculated andmonitored data except at settlement plate P-013 andP-020 where there is underestimation of surface
settlement. However, in general the calculated
312

GEOMATE- Bangkok, Nov. 14-16, 2016

Geomechanics and Geoengineering, 2010.
5(4): p. 223-236.
[3] Bo, M.W., et al., Soil Improvement:
Prefabricated Vertical Drain Techniques.
2003: Thomson Learning, Singapore.
[4] Saowapakpiboon, J., D.T. Bergado, and S.
Youwai, Measured and predicted performanceof prefabricated vertical drains (PVDs) withand without vacuum preloading. Geotextiles
and Geomembranes, 2010. 28(1): p. 1-11.
[5] Mesri, G. and A.Q. Khan, Ground
improvement using vacuum loading together
with vertical drains. Journal of Geotechnicaland Geoenvironmental Engineering, 2012.
138(6): p. 680-689.
[6] Chai, J., S. Hayashi, and J.P. Carter, Vacuum
consolidation and its combination withembankement
loading,inGroundModification and Seismic Mitigation - GSP
152, ASCE, Editor. 2006.
[7] Chai, J., Z. Hong, and S. Shen, Vacuum-drain
consolidation induced pressure distribution
and ground deformation. Geotextiles andGeomembranes, 2010. 28: p. 525-535.
[8] Indraratna, B., R. Cholachat, and I.
Sathananthan, Analytical and numericalsolutions for a single vertical drain including
the effects of vacuum preloading. Canadian
Geotechincal Journal, 2005. 42: p. 994-1014.
[9] Tran, T.A. and T. Mitachi, Equivalent planestrain modeling of veritcal drains in soft
ground under embankment combined withvacuumpreloading.
ComputersandGeotechnics, 2008. 35: p. 655-672.
[10] Qiu, Q.C., H.H. Mo, and Z.L. Dong, Vacuum
pressure distribution and pore water pressure
variation in the ground improved by vacuum
preloading. Canadian Geotechincal Journal,
2007. 44: p. 1433-1445.
[11] Nguyen, C.O. and T.T. Tran, Consolidation
analysis of Vietnam soft marine clay by finite
difference method with application of constantrate of strain consolidation test, in Fifth
International Conference on Geotechnique,
Construction Materials and Environment, H.
Zakaria, Editor. 2015, The GEOMATEInternational Society: Osaka, Japan. p. 271276.
[12] Bui, T.M., Initial Evaluation of Consolidation
Characteristics of Mekong Soft Clay and
Their Use in Engineering Practice, in Hanoi
Engineering Geology Workshop. 2003: Ha
Noi. p. 1-13.
[13] Takemura, J., Y. Watabe, and M. Tanaka,
Characterization of alluvial deposits in
Mekong Delta, in Characterisation andEngineering Properties of Natural Soils. 2007,
Taylor & Francis Group, London. p. 18051829.

results at settlement plates are best fit with the field
monitoring data. The settlement at plates ranges
from 188 cm to 213 cm from plates P-013 to P-020.
The difference in magnitude of settlement can be
designated to the discrepancy of soft clay thickness
over the installation location of the plates.
The compressibility of the soft clay deposit
determined from CRS testing results show good
agreement in magnitude of settlement of the sublayers of soft soil as indicated in Fig. 8. This proves
the appropriateness of the strain rate of 0.02%/min.
used in CRS tests to obtain consolidation
characteristic of the soft soil.
CONCLUSION
There are the following conclusions
Pre-overburdened pressure of the consideredsoft soil deposit is 20 kPa. The soft clay deposit
is in lightly over consolidated state.
2. Compressibility and coefficient of consolidation
derived from constant rate of strain (CRS)
consolidation test at 0.02%/min. can be utilized
directly into the consolidation analysis whose
results consistent with those from field
monitoring.
3. Horizontal coefficient of consolidation is as
much as three times of vertical value determinedby CRS test results c h =3c v . The verticalcoefficient of consolidation in over-consolidated
stage is almost 10 times as much as that innormally consolidated stage c v(OC) =10c v(NC) .
4. The soft ground should be divided into vacuum
zone which extends to the maximum depth of
10 m from the elevation of vacuum pump,which follows Pascal’s law. The ground is
improvement much on the top 10 m, while thebelow part is not effectively improved because
the vacuum pressure cannot be transmitted
effectively deeper than 10 m from vacuum
pump elevation.
1.

REFERENCES
[1]

[2]

Kjellman, W., Consolidation of clayey soils bymeansofatmosphericpressure,
inProceedings of the conference on soil
stabilization, MIT, Editor. 1952: Boston. p.
258-263.
Indraratna,
B.,
X.
Geng,
and
C.
Rujikiatkamjorn, Review of methods ofanalysis for the use of vacuum preloading andvertical drains for soft clay improvement.

313

SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EARTHQUAKE ATTENUATION MODELS AND ITS RESPONSES TO
EARTH ZONE DAM IN UPPER NORTHERN PART OF THAILAND
Tawatchai Tanchaisawat1*, Sirikanya Laosuwan2 and Nutapong Hirano3*Lecturer (tawatchai@eng.cmu.ac.th), 2Doctoral Student, 3Master Student
Department of Civil Engineering, Faculty of Engineering, Chiang Mai University,
Chiang Mai, THAILAND
1

ABSTRACTThe upper northern part of Thailand is mountainous area which classified as moderate risk by earthquake.
This study was focused on the earthquake attenuation model of Abrahamson and Silva (2008). Exciting quake
source from the Chiang Rai 2014 earthquake was recorded at earth zone dam site study. The accelerometer was
installed in the bottom of dam which is 59 m height, 9 m width and 1,950 m length. The attenuation model is therelationship between acceleration, distance, size of earthquake, period of stabilization, characteristics of seismic
sources and geological conditions. This research aims to compare the peak ground acceleration (PGA) between
the attenuation models and the data measurement at the dam to check the accuracy of the model. And to study
the dynamics analysis in the dam by using two selected time history records with different wave characteristics
as: short period (EQ1) and long period (EQ2) to analyze the behavior of the pore water pressure, the stress anddisplacement in the dam. The results of this study revealed that the Peak Ground Acceleration (PGA) from theattenuation model of Abrahamson and Silva equal to 0.017787 g. And the percentage error of calculated PGA
when compared to the PGA from measurement was 6.38. The analysis of the dynamics by using seismic waves
with different time duration. EQ1 have made the pore water pressure on the dam higher when the effective stressin the dam decrease, which makes the displacement of the dam more than EQ2.
Keywords: Earthquake Attenuation Model, Finite Elements Method, Dynamic Analysis, Dam Behavior

INTRODUCTION
Earthquakes are natural disasters that cause
direct damage to engineered constructions such asbuildings, roads, bridges, and dams, causing loss oflife and property. And in the northern part ofThailand has many active faults that causing
earthquake. From seismic records, in the North parthas more frequent earthquakes than the other part.
But currently, Thailand still does not have muchresearch on the decreasing energy of seismic that
depends on the distance. In the past, Warnitchai [1]studied the attenuation model of foreign soil
compared with Bangkok soil. And Attapon [2]studied for the attenuation model, but also focus on
the areas in Chiang Mai.
When the earthquake occurred. The building
structure had effected, particularly the building islocated close to the epicenter of the earthquake. So
this research aims to study the attenuation model byusing the Chiang Rai (6.3 on Richter scale)
earthquake that has epicenter in Phan district,
Chiang Rai province (Latitude 19.748 ºN Longitude
99.692 ºE) on May 5, 2014. The data has collectedby Seismological Bureau of Thai Meteorological
Department as shown in Fig. 1.

Fig. 1 The Chiang Rai earthquake datas (Main shock
and aftershock) have collcted by Seismological
Bureau, TMD
The Abahemsan equation is the attenuation
model that has been used in the research. The
calculated PGA has been compared with the dam
instrumentation (accelerometer) that is installed in
the Mea Ngud Somboon Chon Dam. After that, thedam model will be used to analysis the dam
behaviors (pore pressure, stress and displacement)
after effected by using the earthquake waves thathave short-period and long-period waves. The
Hardening-Soil Model will be used in this analysis.

314

GEOMATE- Bangkok, Nov. 14-16, 2016

When 𝛾𝛾= Unit weight (𝑘𝑘𝑘𝑘/𝑚𝑚2 ), 𝐸𝐸 = Young’s
modulus(𝑘𝑘𝑘𝑘/𝑚𝑚2 ), v = Poisson’s ratio, φ = Friction
angle (°) , c =Cohesion (𝑘𝑘𝑘𝑘/𝑚𝑚2 ) ,ψ = Dilatancy angle (°)

SITE DESCRIPTION AND INSTALLATION
Mae Ngat Dam
Mae Ngat Dam was an earth-filled embankment
dam with a clay core. It is a zoned earth dam
consisting of 4 parts: (1) Core zone, (2) Random
Zone, (3) Miscellaneous, (4) Filter zone and (5)Rock fill zone. The dam was built to block the Mae
Ngat waterway at Cho lae sub district, Mae Taeng
district, Chiang Mai. The dam’s height is 59 meters
and the length is 1,590 meters. The types of soils
used in the construction of Mae Ngat Dam were CL
and MH. The dam’s construction began in 1977 andwas completed in 1984. The cross section of themain dam and its layout is shown in Fig. 2.

Table 2. Properties of soil in Mae Ngat Dam(continue)
Zone
Clay Core
RandomMiscellaneous
FilterPermeability(cm/sec)
0.13 x 10-6
0.96 x 10-6
0.94 x 10-6
1.3 x 10-4

Permeability
Ratio
1
1
1
1

Instrumentation
With the cooperation of Royal Irrigation
Department,
thedaminstrumentations(accelerometers) were installed to measure theacceleration at Mae Ngat Dam in January 2014 in
the middle, on the crest of the dam and the mountainnext to it as shown in Fig. 3.

Fig. 2

Typical cross section of the Mae Ngat Dam.

The Dam’s Materials and Properties
The data of the zoned-earth dam in this study
was collected to acquire detailed information of boththe construction and the timing of water retention by
testing the qualities of the materials both on the field
and in the laboratory by Vikrom [3] and collecting
the samples of soil in the core zone for additional
laboratory testing. The acquired information was
analyzed to select representatives of the current
materials to be able to set properties in creating amodel that is the closest to the current soil
conditions, so it can be used to analyze the behaviorof the dam when affected by Dynamic force.
Properties of materials of the dam are displayed inTable 1 and Table 2.

Fig. 3

Position of Accelerometer installation

The accelerometer model is the 130-SMA(Standard) (Fig. 4). The collected data from installed
instrument can be output in Acceleration-Timegraph. The Peak Ground Acceleration (PGA) can beanalyzed at time period = 0.01. For the Chiang Rai
earthquake (magnitude 6.3 on Richter scale), PGA
can be measured equal to 0.187 m/s2 or 0.019g (Fig.
5).

Table 1. Properties of soil in Mae Ngat Dam
µ
γ
E
Ψ
C φ
Zone
Clay Core 8,171
Random9,620
Miscellane
22,750ous
Filter
9,620

49
32

16 0.48
30 0.350

19.72
18.56

0
0

0

42

0.3

19.0

5

18

28 0.350

15.5

5Fig. 4315

The accelerometer modelGEOMATE- Bangkok, Nov. 14-16, 2016

So the Abrahamson and Silva equation shows in
Eq. (1).
ln( PGA ) = fbase + f fault + AS + f site + fTOR + f dist (1)The 1st function is the variant of the magnitudefand distance term ( base ), as shown in Eq. (2).
acceleration-time
Fig. 5 The
accelerometer.

graphs

from

(

)

(

)

a1 + a4 M - c1 + a8 (8.5 - M )2 +  a2 + a3 M - c1  ln( Rrup 2 + c42 )From the earthquake (magnitude 6.3 on the
Richter scale), which has a distance from theepicenter to Mea Ngud Somboon Chon Dam about
110 kilometers, shown in Fig. 5. The acceleration in
the dam can be measured while the earthquake acted
to the dam. The acceleration-time data records areshown in Fig. 6.

(2)

The 2nd function is the pattern of impact fault
fand aftershocks ( fault + AS ) as Eq. (3).

a12 FRV + a13 FNM + a15 FAS(3)

The 3rd function is the soil characteristics thatfset ( site ), shown in Eq. 4.
V
(a10 + bn) ln  s30 V LIN (4)

The 4th function is the effect of depth to thefmoving plane ( TOR ), as Eq. 5.

Fig. 6 the distance between the Epicenter to Mea
Ngud Somboon Chon Dam
For this event. The analysis results from Global
CMT Catalog conclude that the type of fault is strike
slip and show the parameter of strike, dip, slip andmoment tensor, as shown in Fig. 7.

Z-2)
a16 FRV + a17 (1- FRV )( TOR
8(5)

The 5th function is the affected for small tofmedium earthquake at long distances ( dist ),shown in Eq. 6.
a18 ( Rrup -100)(6.5 - M )Fig. 7 Analysis of earthquake mechanisms from
Global CMT Catalog [4].

(6)

When M=Moment magnitude, R rup =Rupture
distance (km), R jb =Joyner-Boore distance (km),R x =Horizontal distance (km) from top edge ofrupture, Z top =Depth-to -top of rupture (km),F RV =Flag for reverse faulting earthquakes,
F NM =Flag for normal faulting earthquakes,
F AS =Flag for aftershocks, F HW =Flag for hanging
wall sites, Dip=Fault dip in degrees, V S30 =Shearwave velocity over the top 30 m (m/s) Z 1.0 =Depth toV S =1.0 km/s at the site (m), PĜA 1100 =Median peakacceleration (g) for V S30 =1100 m/s and W=Downdip rupture width (km). The value for the PeakGround Acceleration (PGA) that comes from thedefinition of Abrahamson and Silva (2008)

THEORYIn this research was conducted to study theappropriate model for attenuation equation in thenorthern part of Thailand. The previous research,
Nutapong [5] studied the attenuation model andconcluded that the equations of Abrahamson andSilva [6], the one of the NGA model developed fromthe Abrahamson and Silva model (1997), is suitablefor use to analyze the attenuation model in thenorthern part of Thailand.

316

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 3. The coefficient of Peak Ground Acceleration (PGA) [4]
c1
c4
a3
a4
n
V LIN

b

a1

a2

a8

0.725

-0.968

0

6.75

4.5

0.265

-0.231

1.18

865.1

-1.186

a 10

a 12

V S30

a 13

a 15

a 16

a 17

a 18

0.9485

-0.12

1100

-0.05

-0.4050.65

0.6-0.0067

Calculation of the PGA of an earthquake in
Chiang Rai.

set to Hydrostatic Pressure. Then, the experiment
was analyzed under Dynamic force.

fbase = −4.28256

f fault + AS = 0

f site = −0.10833
fTOR = 0.375
f dist = −0.0134

ln( PGA) = (−4.28256) + (0) + (−0.10833) + (0.375) + (−0.0134)
PGA = 0.017787

Fig. 8 Triangular elements with 15 Nodes and 12
Stress points.

RESULTS
Attenuation Model
From using the Abrahamson and Silva equationin the attenuation model analysis, the PGA from
Chiang Rai earthquake at Mea Ngud Somboon Choncan be resolve. The parameters that were used in thecalculation consists of M is equal to 6.3 on the
Richter scale, ZTOR equal to the depth of 7 km,
RRUP than 110 km from the spot. The Chiang Rai
earthquake occurred at Payao fault which ishorizontal movement (Strike slip). The result foundthat the Peak Ground Acceleration (PGA) from
calculation equal to 0.017787 g. And the percentage
error of calculated PGA when compared to the PGA
from measurement was 6.38.

Fig. 9 Mae Ngat Dam model using finite element
method (FEM) consists of 845 elements and 6947
nodes.

Fig. 10 The model using finite element method
shows the Phreatic line of Mae Ngat Dam.

Dynamic Analysis

The 2 difference earthquake waves (short-time
period; EQ1 and long-time period; EQ2) (Table 4)
was used to analyze for study the behavior of porewater pressure, stress and displacement that occurs
in the dam during earthquake occurs. The excess
pore water pressure was considered at 37 m depth
from the dam crest. The results were shown in Fig.
11-18.

The data used in the analysis came from the soils
used in the dam construction which were tested in
the field and the laboratory. The datas were used tocreate the model dam by using the Hardening-Soilmodel (2D shear plane) with the Finite Element
Method to study the response of the dam underearthquake loads. The PLAXIS program [7] wasused to analyze the model. The selected elements
were triangular with 15 Nodes and 12 Stress points(Fig. 8) from Brinkgreve and Vermeer [8]. The
water level in the dam was analyzed while the water
retention level was normal. The water pressure was317

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 4 The data of seismic waves used in theanalysis.
Earthquake
Victoria
BorregoMexico (EQ1) Mth. (EQ2)
Year (A.C.)
1980
1968
PGA (g)0.130
0.120Time (sec)
15.58
40
Magnitude (M)
6.33
6.63
Distance (km.)
58.87
70.75
Predominant
0.30
0.33
periodType of
Short periodLong period
earthquake

Fig. 13 The excess pore water pressure occurring in
the dam under short period earthquake (EQ1).

The study found that the pore water pressure in
the dam was increasing during the seismic force
acted to the dam. The long period of earthquake(EQ2) make the pore water pressure is worth morethan the short period of earthquake (EQ1) when the
PGA of two difference waves are similar. In Fig. 1112, the range of changing in pore water pressure of
approximately 102.5-107.5 kPa under EQ1 and116.75-121.5 kPa under EQ2. The excess pore waterpressure occurring in the dam of EQ1 is equal to66.48 kN/m2 and EQ2 is equal to 89.17 kN/m2,
respectively in Fig. 13-14.

Fig. 14 The excess pore water pressure occurring in
the dam under long period earthquake (EQ2).

Fig. 15 The effective stress of the dam under shortperiod earthquake (EQ1).
Fig. 11 The pore water pressure changing undershort period earthquake (EQ1).

Fig. 16 The effective stress of the dam under long
period earthquake (EQ2).

Fig. 12 The pore water pressure changing underlong period earthquake (EQ2).

318

GEOMATE- Bangkok, Nov. 14-16, 2016

wave (EQ1). Under the seismic waves that have the
PGA about 0.12-0.13g, they make the excess porewater pressure the pore water pressure increases
about 12-15%, the maximum effective stress is equal
to 1980 kN/m2 and 1080 kN/m2 for EQ1 and EQ2
respectively. When the pore water pressure andstress in the dam were changed by the seismic force,
they are important factors that cause thedisplacement of the dam and cause a decrease in thestrength of the soil. So the total displacement can beanalyzed equal to 43.01 mm. and 23.23 mm. for
EQ1 and EQ2 respectively.

Fig. 17 The total displacements of the dam undershort period earthquake (EQ1).

ACKNOWLEDGEMENTS
Thanks to all the professors in the civilengineering department who provided fundamental
knowledge on geotechnology and thank you to the
Mae Ngat Dam Royal Irrigation Department, The
Ministry of Agriculture and Cooperatives, theConstruction Management Department, and theRegional Irrigation Office 1 Chiang Mai, Daminstrument Ltd., Part., who was the supervisor, andfor the information and convenience provided while
conducting field work for this study.

Fig. 18 The total displacements of the dam undershort period earthquake (EQ2).

REFERENCES

From the seismic force, not only caused the porewater pressure in the dam increase but it also affects
to the effective stress. The maximum effective stressthat could be calculated at core zone at the base of
the dam during the earthquake was 1980 kN/m2 for
EQ1 and 1080 kN/m2 for EQ2. In Fig. 15-16 show
the shade area to present the effective stress in theearth zone dam.
The maximum total displacement equal to 23.23mm and 43.01 mm for EQ1 and EQ2 condition
respectively. The displacement from theseearthquakes are shown in Fig. 17-18. The results
showed that the maximum total displacement of themodel affected by the short period of earthquake
occurred in the downstream side of random zone andchanged to upstream side when it affected by thelong period of earthquake.

[1] Warnitchai P, Sangarayakul C and Ashford S,“Seismic Hazard in Bangkok Due to LongDistance Earthquake”, Proceedings of the 12th
World Conference on Earthquake Engineering,
Auckland, New Zealand, 2000
[2] Attapon A, “Earthquake attenuation model for
Chiang Mai subsoil”, Chiang Mai University,
Thailand, 2011.
[3] Vigrom V, “The behavior of Mae Gnad earthfill
dam during construction”, Kasetsart University,
Thailand 1987.
[4] Global CMT Catalog Search, Available
at: http://www.globalcmt.org/CMTsearch.html,
2016[5] Nutapong H, “Analysis of the earthquake
attenuation model to the 2014 Chiang Rai
earthquake affected to Mae Ngud Somboon
Chon dam in Chiang Mai”, The 21st National
Convention on Civil Engineering, 28-30 June
2016, Songkhla, Thailand.
[6] Abrahamson N.A. and Silva W.J., “Summary of
the Abrahamson & Silva NGA Ground Motion
Relations”. Earthquake Spectra. 24, 1: 67–97,
2008[7] PLAXIS 2D version 8.2, Version 8.0 Plaxis BV,
Delft.
[8] R.B.J Brinkgreve and P.A. Vermeer., “Finite
element code for soil and rock Analysis”,
PLAXIS 8.0 manual, Balkema, Rotterdam, The
Netherlands, 1998.

CONCLUSION
After comparing the analysis results with themeasurement value from the accelerometer foundthat the attenuation model of Silva and Abrahamson(2008) can calculated the Peak Ground Acceleration(PGA) equal to 0.017787 g and the percentage error
was 6.87.
From study the pore water pressure in the zone
earth dam in dynamic condition by using finite
element method found that the pore water pressure
in dam under the long period earthquake wave
(EQ2) is greater than the short period earthquake319

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

VIBRATION CHARACTERISTICS OF DEFORMED STONE WALLS
OF JAPANESE TRADITIONAL CASTLE
Minoru Yamanaka1, Hayato Ishigaki2, Katsuhiko Koizumi2, Shuichi Hasegawa1 and Hiroyuki Araki3
1

Department of Safety System Construction Engineering, Kagawa University, Japan;
2
Graduate School of Engineering, Kagawa University, Japan;
3
Department of civil and Environmental Engineering, Chuo University, Japan
ABSTRACTThe non-destructive survey technique is required to estimate quantitatively the deterioration level of a stone
wall at a traditional castle site. Microtremor measurement method is one of the non-destructive survey techniqueto estimate geotechnical characteristics of the ground. In this study, microtremors are measured at the stone wallof the Marugame castle, in order to evaluate vibration characteristics of deformed or non-deformed stone wall; themethod to extract the deformation part of the stone wall is discussed based on the measurement results. The results
showed that the value of H/V and H/H spectral ration increased with the increase of deformation of wall but also
by the elevation of a measurement point. It was observed that the values of energy ratio were scattered widely. It
is concluded that the H/H spectral ratio is a suitable index to indicated the deformation level of a stone wall. By
setting the adequate threshold value, deformed walls were extracted roughly by the value of H/H spectral ratio at
the Marugame castle.
Keywords: Microtremor measurement, stone wall, H/V spectral ratio, H/H spectral ratio, Japanese castle
measuring target is estimated approximately based on
the ratio of horizontal and vertical spectrum of the
microtremor, H/V spectral ratio.
A few study on application of microtremor
measurement method for evaluation of vibration
characteristics of stone wall has been conducted,
while the microtremor measurement method wassometimes applied to the evaluation of the ground andstructures. For example, the microtremor
measurement method was applied to the stone wall at
the Kochi castle. The measurement results explainedthat the H/V spectral ratio was one of the deteriorationindex of the stone wall [4]. However, the microtremorof stone walls with highly deformation has not been

INTRODUCTION
Japanese traditional castles have been often
constructed at a crest of soil embankment. The wall
face of the soil embankment was commonly protectedwith stones. These stone walls should be left for
posterity as a part of cultural heritage.
However, stone walls deteriorated with ageing are
affected sometimes seismic damage. For example, thestone walls of Sendai castle ruin and the KomineShirakawa castle ruin was collapsed by The 2011 off
the Pacific coast of Tohoku Earthquake. Recently,
stone wall of the Kumamoto castle was collapsed by
The 2016 Kumamoto Earthquake.
In order to keep for posterity, a stone wall should
be maintained according to its deterioration level. The
deterioration level of a stone wall is ordinary judged
based on visual observations in appearance condition.
In-situ tests such as boring and sounding test havebeen hardly used in survey of stone walls to avoid
disturbance of a wall [1]. Especially the influence ofin-situ tests for a stone wall which is collected in thecultural asset must be minimized from the viewpointof cultural property protection. Therefore, the nondestructive survey technique is required to estimate
quantitatively the deterioration level of a stone wall.
Microtremor measurement method is one of thenon-destructive survey technique to evaluate
vibration characteristics of the ground and structures[2],[3]. A predominant period is evaluated fromspectrum of microtremor. S-wave response of theFig. 1320

The layout of the Marugame Castle
plotted on geological map.

GEOMATE- Bangkok, Nov. 14-16, 2016a) Swelling and opening (Sannomaru)b) Habaki stone wall (south-west part of
Honmaru)

Fig. 2 Typical deformed stone wall a) and Habaki stone wall b)measured in the previous study.
In this study, microtremor measurement method
is applied to the stone walls with highly deformation
at the Marugame castle, Kagawa prefecture in Japan.
The method to extract the deformation part of thestone wall is discussed based on the measurement
results
OUTLINE OF MARUGAME CASTLE SITE
Figure 1 is the layout of the Marugame castle
plotted on geological map [5]. The Marugame castle
built in 1602, is located on the crest of Mt. Kameyama
consisting of andesite and granite. The Marugame
castle is composed of Honmaru (donjon), Ninomaru(first outer citadel), San-nomaru (second outercitadel) and Obikuruwa (outermost citadel). Eachcitadel is supported by a stone wall. The height of thewhole stone wall is tallest in that of Japanese castle.
Figure 2 shows typical deformation of stone wall
observed at the Marugame castle. Figure 2a showsswelling of a wall face and opening of the joint
between stones. The reason of these deformations are
estimated to earth pressure and rain water seepage.
‘Habaki’ stone wall was built at the foot of stone walls
in order to restrain the deformation of the walls, as
shown in Fig. 2b.
Figure 3 shows a schematic of a cross section of
typical stone wall at the Marugame castle. A stone
wall composed of a wall face stone, a crushed stone
zone, backfill and bed rock. The width of the crushedstone zone is assumed to be 1 to 2 m approximately.
MICROTREMORSTONE WALL

MEASUREMENT

OF

Fig. 3 Schematic of a cross section of a typicalstone wall at the Marugame castle
and a toe of a wall behind in order to ensure to bereference point.
Figure 4 shows the planform of microtremor
meters location. The numbers of MWS, MCS andMBR are 133, 131 and 9, respectively. The distancesof each MWS and each MCS were 5 m approximately.
The data of microtremor were measured at MWS,
MCS and MBR at the same time. The measurement
time was about 1 hour to avoid effects of vibration ofpedestrians.
Microtremors were measured at the geological cliffof the Sendai castle in previous study [6], whichshowed that the ground motion amplification of theperpendicular direction to the cliff was larger thanthat of the horizontal direction. Referring to theprevious study, N-S direction of a microtremor meter
was set to the perpendicular direction to a stone wallin this study, N-S component of microtremors wasanalyzed mainly in order to observe the motion inperpendicular direction to a stone wall.

A

The location of microtremor meters, McSEIS-MT
NEO made by OYO corporation in Japan, are also
shown in Figure 3. Microtremor meters were put
directly on a wall face stone, a crushed stone and bed
rock, which were named as MWS, MCS and MBR in
this paper, respectively. The distance between MWS
and MCS was 2 m approximately. MBR weretypically put on the midpoint between a top of a wall

MICROTREMOR MEASUREMENT RESULTS
AND DISCUSSION
Analysis method of microtremor records
321

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 4The location of microtremor meters at the Marugame castle

Microtremor records were used to analysis thevibration characteristics of Marugame castle. All
microtremor spectrum (Fourier amplitude spectrum)
were obtained in a frequency range of 1 to 10 Hz, thenthose were smoothed separately with Parzen window
with a band width of 0.3 Hz. In order to found thenormalized result explaining clearly the relationships
between vibration characteristic and deformation ofstone wall, three kinds of normalized result of
microtremor records, as H/V spectral ratio, H/H
spectral ratio and energy ratio were calculated basedon the microtremor records.
H/V spectral ratio and H/H spectral ratio
H/V spectrum is a ratio of horizontal component
and vertical component of a Fourier amplitude
spectrum in perpendicular direction to a stone wall
(NS component of a microtremor). The H/V spectral
ratio is the H/V spectrum obtained at MWS or MCS
normalized by the H/V spectrum obtained at MBR.
The H/H spectral ratio, defined in this study, ishorizontal component of a Fourier amplitude
spectrum obtained at MWS or MCS normalized byhorizontal component of a Fourier amplitude
spectrum obtained at MBR in perpendicular direction
to a stone wall.
Figure 5 shows typical spectral ratios of H/V andH/H obtained at 2 sites where the stone walls were
judged to be largely deformed (W3-27) and beslightly deformed (W3-10) based on the visualobservation in appearance condition. Black circles
shown in Figure 5 indicate the peak values of thespectra. The peak values of spectral ratio wereobtained in all observation points in the same manner.

Fig. 5322

Typical H/V and H/H spectral ratios
obtained at W3-27 a) and W3-10 b)

GEOMATE- Bangkok, Nov. 14-16, 2016

Energy ratio
Yamauchi et al. [7] pointed out that the value of a
Fourier amplitude which was obtained at unstablerock increased in all frequency range before the rockslid down. In order to clarify the increase of theamplitude values which were obtained at unstablestone of a stone wall, the value of energy was
obtained by integration of Fourier amplitude spectraof 3 components, vertical and 2 horizontal
components. Then the value of energy at MWS or
MCS was normalized by the value of energy at MBR,
which was defined as energy ratio in this paper.
Comparison of spectral ratios and energy ratio
The results of H/V spectral ratio and H/H spectral
ratio are shown in Fig.6 and Fig. 7, respectively. The
results measured at highly deformed site and small
deformation site were plotted with red legend andblue legend, respectively.
The value of H/V spectral ratio obtained at highly
deformed site was slightly larger than that at small
deformation site, while the value scattered widely(Fig.6). A similar trend was observed also in the
results of H/H spectral ratio (Fig.7). The variation of
the value of H/H spectral ratio was smaller than thatof the value of H/V spectral ratio (Fig.6 and 7). The
values of H/V and H/H spectral ratio increased with
the increase of deformation.
In addition, the value of H/V spectral ratio
obtained at Honmaru was larger than those at
Ninomaru and Sannomaru (Fig.6). The elevation of
Honmaru is higher than those of Ninomaru andSannomaru. Hence, the value of H/V spectral ratio
was affected possibly by the effect of not only
deformation but elevation of a measurement point.
On the other hand, the results of H/H spectral ratio
were less affected by elevation of a measurement
point. Therefore, H/H spectral ratio is assumed to be
one of suitable index to extract highly deformed site.
The values of energy ratio obtained at allmeasurement points are shown in Fig. 9. The value ofenergy ratio obtained at MWS varies widely and is
totally larger than that obtained at MCS. This trend
explains the amplitude of micromotion at MWS islarger than that at MCS. The value of MWS which isremarkably large compared with the value of MCS
might suggest existing of unstable stone in the stone
wall. However, it is difficult to use the value of energy
ratio for extraction of a deformed part of a stone wall.

Fig. 6

H/V spectral ratio obtained at MWS and
MCS

Fig. 7

H/H spectral ratio obtained at MWS and
MCS

Fig. 8Energy ratio obtained at MWS and MCS

Extracting method of a deformed stone wall byusing H/H spectral ratio
As described above, the value of H/H spectral
ratio was assumed to be one of an index to extract
highly deformed site. The value of H/H spectral ratio
was compared with the deformation state of a stone
wall in order to conform the availability of this value.

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GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 9The location of deformed wall and the distribution of the thresholded value of H/H ratio

Figure 9 shows the location of deformed stone
wall. The 60% of the whole stone walls was judged
as highly deformed by visual observations.
Based on this rate, upper 60 % of the value of H/H
spectral ratio was assumed to be obtained at thehighly deformed wall. The lower threshold value of
H/H spectral ratio including upper 60 % was obtained
as 4.0 in both cases of MWS and MCS, then thethreshold value was set to 4.0 to extract highly
deformed wall. Figure 9 also shows the distributionof the threshold value of H/H spectral ratio. The
measurement points were classified into 3 types, both
values at MWS and MCS exceeding 4.0 (red circle),
one value at MWS or MCS exceeding 4.0 (yellow
circle) and both values at MWS or MCS fallingbehind 4.0 (blue circle). The number of red circle,
yellow circle and blue circle was 78, 16 and 32,
respectively.
The deformed walls were observed at 47 sites in
78 red circles, 6 sites in 16 yellow circles and 2 sites
in 32 red circles. When the both value of H/H spectral
ratio at MWS and MCS exceed 4.0, the stone wallunder the measurement point was highly deformed
with a probability of about 60 %. This result suggest
the deformed walls are extracted roughly by the valueof H/H spectral ratio and adequate threshold value.
Further study is required to reveal the setting methodof threshold value.

stone wall of Marugame castle in order to evaluate
vibration characteristics of deformed or nondeformed stone wall. The following conclusions
could be derived from the results of the microtremor
measurement.
1) The values of H/V spectral ratio scatteredwidely and slightly increased with the increase of thedeformation. The value of H/V spectral ratio was also
affected possibly by the effect of the elevation of ameasurement point.
2) The values of H/H spectral ratio increased with
the increase of deformation. In addition, the values of
H/H spectral ratio were less affected by the elevationof a measurement point.
3) The H/H spectral ratio is a suitable index toindicate the deformation level of a stone wall. The
deformed walls are extracted by the value of H/H
spectral ratio and adequate threshold value, 4.0 at the
Marugame castle. When the both values of H/H
spectral ratio at MWS and MCS exceed 4.0, the stone
wall might be highly deformed with a probability of
about 60 %.
4) It is difficult to use the value of energy ratio for
extraction of a deformed part of a stone wall because
the value scattered widely.
REFERENCES[1] Nishida K., Ohshima K. Kanaoka M. andKitazono K. “Geotechnical survey on the stone
wall at Takamatsu castle”, in Proceedings of the
46th Annual Conference of the Japanese Society

CONCLUSIONSIn this study, microtremors were measured at the324

GEOMATE- Bangkok, Nov. 14-16, 2016

Geotechnical Society, 2015, pp. 203–204 (in Japanese).

of Civil Engineering, No. 4, 2009, pp.171–172.
(in Japanese)[2] Yamanaka M., Hachimura T. and Hasegawa S.
“Estimation of thickness of landfill by
geophysical exploration methods at illegal
industrial wastes disposal site”, Proceedings of
4th International Conference on Geotechnique,
Construction Materials and Environment;
GEOMATE2014, Brisbane, pp.232-237, 2014.
[3] Putra R.R., Kiyono J., Yoshimoto Y., One Y. and
Syahril, “Determined soil characteristic of Paluin Indonesia by using microtremor observation”,
International Journal of GEOMATE, Vol.10,
No.2, pp.1737-1742, 2016.
[4] Yamanaka M., Nishiwaki D., Koizumi K. and[5] National Institute of Advanced Industrial Science andTechnology, Geological Survey of Japan, “Geologicalmap
displaysystemGeomapNavi”,
https://gbank.gsj.jp/geonavi/geonavi.php#latlon/16,34.28567,133.80458[6] Motosaka M. and Watanabe T. “Ground vibration
characteristics of geological cliff structure at the
Sendai castle site based on microtremor andearthquake observations and prediction of strong
ground motion spectra for past earthquake”, Journal of
Structural Engineering B, Architectural Society ofJapan, Vol.47B, 2001, pp.167–172. (in Japanese)[7] Yamauchi Y., Jiang Y. and Tanabashi Y. “Monitoring
management of the unstable rock blocks aimed at thechanges of a dynamic characteristic”, Proceedings ofsymposium on geological environment andgeotechnical survey, Japanese Geotechnical Society,2006, pp.21–26. (in Japanese)

Hasegawa S. “Detection of dynamic unstable point ofstone wall at Kochi Castle by microtremor”,
Proceedings of 50th Annual Conference of Japanese325

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DETERMINATION OF THE POST-CYCLIC YIELD STRENGTH AND
INITIAL STIFFNESS OF TWO PEAT SOILS

1

Adnan Zainorabidin1, Habib Musa Mohamad2
Faculty of Civil and Environmental Engineering, Universiti Tun Hussein Onn Malaysia, Malaysia2

ABSTRACTThe post-cyclic yield shear strength and initial stiffness of a peat soil after subjecting to cyclic loading is amajor topic in this study. Due to the effects of cyclic loading, post-cyclic shear strength decreases lower than its
initial strength. A series laboratory static and cyclic triaxial test followed by post-cyclic monotonic tests carriedout to determine the yielding parameters. Tests were carried out on the undisturbed samples taken from Parit
Nipah, Johor and Lumadan, Sabah within west and east Malaysia peat soils. Post-cyclic loading test conductedwith effective stress 100 kPa with frequency 1.0Hz. The initial stiffness is the initial tangential modulus with yield
shear strength was half of the deviator stress in which two tangential lines intersect. The post cyclic yield shear
strength of undisturbed peat soil is considerably lower compared to static at the axial strain of only 1.4% and 1.5%.
The Parit Nipah and Lumadan peat are classified as Hemic.
Keywords: Post-cyclic, Peat, Yield shear strength, Initial stiffness

Post-cyclic behaviour of soil is generally
considered to depend on the maximum strain
developed during cyclic loading [8]. Post-cyclic as
tests carried out on the sample by allowing the cyclicpore pressure to develop during cyclic loading toeither dissipate or without dissipation [9]. While thesamples are then subjected to post-cyclic monotonic
loading without dissipation of cyclic pore pressure.
Post-cyclic monotonic shear strengths were evaluatedafter various numbers of cycles of dynamic loading[10].
The studies of post-cyclic behaviour of soils
includes the loss of static undrained shear strengthand strain softening of soils under cyclic loads [11].
There is a considerable reduction in the undrained
strength after cyclic loading [12]. There were effects
from the number of loading cycles on the test
respectively involved. Similar trends were observed
for the test, and it shows that there were increases
with continuing loading cycles from 1 to 100 cycles[9]-[13] findings.
The initial stiffness is the initial tangential
modulus, which is in turn the slope of the curve ofdeviator stress versus axial strain at the axial strain of
0%. To get the yield shear strength, two tangential
lines were plotted [14]. The yield shear strength was
half of the deviator stress at an axial strain, in whichthose two tangential lines intersect [15]. The shear
strength and stiffness at small deformation werecalled as yield shear strength and initial stiffness
while, the shear strength and stiffness at large
deformation were called as undrained shear strengthand secant modulus, respectively.
Figure 1 shows the illustration of method for
determination of yield shear strength (Sy) and initial

INTRODUCTION
Peats occur as extremely soft, wet andunconsolidated surficial deposits.
Peats aregeotechnically problematic due to their highcompressibility and low shear strength [1]. In amoderate load increment, it may lead to large volume
changes that shows high compressibility. Deep peats
exhibit high compressibility, medium to low
permeability, low strength and volume instability [2].
Peats are also characterized by high initial void ratio,organic content and water holding capacity [3].
Since the main component is an organic matter,peats are very spongy, highly compressible andcombustible
incharacteristics
[4].
These
characteristics make the peats to pose its own
distinctive geotechnical properties as compared toother inorganic soils like clay and sandy soils which
are made up of only soil particles [3]. Peats which
are formed from the accumulation of organic
materials over thousands of years, are characterisedby its high water content, compressibility and low
shear stiffness and shear strength [5]. However, theoverburden pressure of peats are very low [6].
Nonetheless, the effective shear strength of peatsoils are essentially a frictional material and that it
behaves closely in accordance with the principles ofeffective stress.
Past researchers conclusively
explained that standard tests that consolidates
undrained triaxial test with the measurement of porewater pressure does not required over 50% axial
strain to fail [6]. The undrained shear strength (su)refers to the strength of soil in situations where theexcess pore water pressures developed during
shearing cannot dissipate and failure takes place [7].
326GEOMATE- Bangkok, Nov. 14-16, 2016

Dynamic triaxial apparatus or popularly known as
Dynamic Triaxial Testing System (ELDYN) wasused in the determination of shear strength of peat soil
using an electronic controlled system. Sample
preparation for the post-cyclic triaxial test is similar
to the monotonic test. The specimens was mounted
on the base of the pedestal sealed with a rubber
membrane and ends with filter paper and porous stone
at each end. All samples were consolidated 100 kPa
effective confining stress and cyclic tests were
performed under 1.0 Hz frequency in order todetermine the shear strength. The index properties
tests conducted on undisturbed specimens.
Monotonic loading was seriate applied at a0.10mm/minute loading rate, and lasted till the soil
specimens exhibited an axial strain of 20%. The
specimens were subjected to 100 cycles. After cyclicloading, the specimens were immediately subjected topost-cyclic loading or known similar to standard
consolidated undrained triaxial monotonic loading to
failure.
Peat sample was obtained from Malaysia peat
deposit area. Under reserved land area and far fromagricultural activity. As seen in Table 1, the index
properties of Parit Nipah fairly significant that natural
moisture content is about 593% higher than Lumadan
peat about 455.51%. The natural water content ofpeat in Malaysia ranges from 200 % to 700 % andwith organic content in the range of 50 % to 95 % [1].
Therefore, the recorded values for Parit Nipah andLumadan fulfil this statement. Specific gravity
recorded 1.3 for Parit Nipah and Lumadan recorded
1.37 were within the range as reported by [1]. In
addition to basic characterization tests, the Parit
Nipah and Lumadan peat identified as Hemic.

stiffness (Ei) suggested by Wang (2011). The initialstiffness is the initial tangential modulus, which is inturn the slope of the curve of deviator stress versus
axial strain at the axial strain of 0%. Two tangential
lines were plotted to get the yield shear strength. The
yield of shear strength was half of the deviator stressat an axial strain and intersect with two tangential
lines [15]. This research work deals with a study on
the post cyclic behaviour of peat soils analyzed using
dynamic and static methods. The application ofeffective stresses and frequencies are criticallyreviewed using triaxial tests.

Fig. 1

Determination of yield shear strength andinitial stiffness (Wang, 2011).

A series of triaxial tests were carried out toinvestigate the influence of dynamic loading on shear
strength. The post-cyclic yield shear strength andinitial stiffness of peat soil after subjecting to cyclicloading is presented in this research. The procedureused to determine the stress-strain threshold is
proposed using the reduction of post-cyclic strength
after 100 cycles of loading.

Table 1 Index properties of peat soil

MATERIAL AND EXPERIMENTAL METHODThis paper had conducted investigation on the
post-cyclic shear strength of peat soil. All tests havebeen conducted using undisturbed peat soil specimen.
The sampling location was located in Parit Nipah,
Johor and Lumadan, Beaufort, Sabah. Ground watertable was found at the depth of less than 1 meter
during the sampling. The soil was excavated to adepth of 0.5 m below the ground surface and numbersof tube sampler with the size of 50mm diameter and100mm height were pushed slowly into the soil. The
undisturbed peat soils were waxed both at the end of
the tubes and sealed with the aluminium and plastics
to prevent the loss or gain of moisture. Jolting duringtransport was avoided.
The samples were kept in the laboratory underconstant temperature in the air conditioned room.

Properties

LumadanParit Nipah

Moisture content, %455.51593

Liquid limit wl, %Specific Gravity, Gs
pH testOrganic Content, %Fiber Content, %Von Post Scale

211
1.37
4.3
95.51
66
H7

243
1.3
4.0
95.6
38.5
H5

DETERMINATION OF YIELD STRENGTH
AND INNITIAL STIFFNESSIn order to evaluate the post-cyclic yield strengthand initial stiffness, an extensive series of static andcyclic tests were carried out on large 50 mm diameter
cylindrical specimens of peat soil materials using
consolidated undrained triaxle tests and followed by327

GEOMATE- Bangkok, Nov. 14-16, 2016

post-cyclic shear test after cyclic loading. This paper
presented the behaviour of peat soil subjected cyclicloading and effect to post-cyclic shear strength while
compared to static results.
The shear strength and stiffness at small
deformation were called as yield shear strength andinitial stiffness, respectively and the shear strengthand stiffness at large deformation were called as
undrained shear strength and secant modulus [14].
This research carried out to study the shear strengthand stiffness at small deformation. Static results arecompared to the post-cyclic shear strength andstiffness. The reduction of shear strength andstiffness measured and there are significant behaviour
when peat imposed with cyclic loading.
Figure 2 shows the consolidated undrained static
results for both sample. Parit Nipah peat has
maximum shear strength at 92.04 kPa and higher than
Lumadan peat where about 58.44 kPa. The variationsof index properties of both sample resulting indifferent shear strength property. The West Malaysia
peat moisture content varied in the ranges 676.30% to735.45% [16] while, 710.44% [17], 460% [18].
Undoubted, tropical peat has high water content andmost of peat deposit are in moderately decomposed
condition. However, the shear strength of intact peat
is made up of interparticle friction as well as tension
in the peat fibres [19].

Fig. 2

Table 2 indicates the initial stiffness and shear
strength (Sy) results for static and post-cyclic test
conducted. In post-cyclic, Lumadan peat shear
strength 21.55 kPa, decreased compared to static
about 29.37 kPa. This reduction of shear strength
Lumadan peat fairly lost in shear strength about
26.63% from the initial strength. Parit Nipah peat incertain patterns keep popping up in same behaviour.
Those relationship offers decreases in shear strength
after subjected to cyclic loading. Initial shear strengthof Parit Nipah peat 46.02 kPa, decreased to 60.13%
and remaining shear strength about 18.35 kPa. The
static undrained shear strength decreased
significantly after 15 loading cycles and softening
behaviour occurred when the specimen exhibited ±1.5% axial strain [10].
The initial stiffness (Ei) of both samples expressly
decrement in post-cyclic compared to static results.
The reduction of initial stiffness of Lumadan andParit Nipah peat about 60.80% and 31.92%,
respectively. The initial stiffness slope tangential getpasses through the curve of deviator stress andpointed 10.45 for Lumadan peat and 10.73 for Parit
Nipah.
Tangential slope for both sample almost closed
where the curves derived from deviator stress in postcyclic notched same patterns. The decrement ofinitial stiffness due to cyclic loading. On the other
hand, if peat soil has experienced cyclic loading, thedeformation of soil structure and alteration of void
will resulted in decreasing in yield shear strength.
The existence of yield strength, where softening
starts, is determined from post-cyclic tests. The levelof post-cyclic yield strain occurred in ±0.62% of axial
strain for Parit Nipah peat and ±0.58% for theundisturbed peat soils. The shear deformations
increase rapidly in post-cyclic, exceeding the curves
after 9%. The reduction properties of initial stiffness
and yield shear strength due to soil softening during
cyclic loading. Softening behaviour occurred after
12 cycles, in this study the softening behaviour
outgrow at 100 number of cycles [10].

Consolidated undrained static results

This study focuses on the shear strength and initialstiffness of peat soil in post-cyclic behaviour. Figure
3 shows the results of consolidated undrained postcyclic for both sample. This test carried out after 100
numbers of cycles in dynamic stage. This test similar
to static test arrangement. From the results, it can beseen clearly that, the shear strength of peatsignificantly decreased when subjected to cyclicloading. This statement in line with [16] where, theshear strength of peat soil decreased after 100 numberof cyclic loading in post-cyclic compared to themonotonic tests and post-cyclic peak shear strength
decreased substantially with frequencies applied.

Fig. 3328

Consolidated undrained post-cyclic results

GEOMATE- Bangkok, Nov. 14-16, 2016[4] Shafiee A., Scott J. B. and Jonathan P. S. (2013).
Laboratory Evaluation of Seismic Failure
Mechanisms of Levees on Peat. Department of
Civil
and
Environmental
Engineering
University of California, Los Angeles. Report
UCLA-SGEL
2013/04
StructuralandGeotechnicalEngineering
Laboratory
Department of Civil & Environmental
Engineering University of California, Los
Angeles September 2013. Ph.D Thesis.Wang S.
(2011). Postcyclic behavior of low-plasticity
silt. A dissertation presented to the faculty of the
Graduate School of the Missouri University of
Science and technology. Doctor Philosophy.
[5] Boylan N. and Long M. (2012). Evaluation ofPeat Strength for Stability Assessments.
Proceedings of the Institution of Civil Engineers,
Geotechnical Engineering. University College
Dublin, Ireland. 166: 1-10.
[6] Gosling D. and Keeton P. (2006). Problems
with Testing Peat for Stability Analysis. Peat
Seminar. Edinburgh, 11 March 2008.
[7] Boylan N., Jennings P. and Long M. (2008).
Peat slope failure in Ireland. Quarterly Journalof Engineering Geology and Hydrogeology
2008;
v.
41;
p.
93-108doi:10.1144/14709236/06-028.
[8] Diaz-Rodriguez J. A., Moreno P. and Guadalupe
S. (2000). Undrained Shear Behaviour of
Mexico City Sediments during and After Cyclic
Loading. 12 WCEE. 2000. 1653.
[9] Rakesh J. P., Nazeeh K. M., Robinson R. G.
(2013). Post-Cyclic Behaviour of Clayey Soil.
Indian Geotechnical Journal. March 2014,
Volume 44, Issue 1, pp 39-48.
[10] Erken A. and M.B.C. Ülker. (2008). The PostCyclic Shear Strength of Fine Grained Soils.
The 14th World Conference on Earthquake
Engineering October 12-17, 2008, Beijing,
China.
[11] Alex G. G. (2014). Shear resistance degradationof lime –cement stabilized soil during cyclicloading. Division of Soil- and Rock Mechanics
Department of Civil, Architectural and the Built
Environment Stockholm 2014. Independent
thesis Advanced level (degree of Master.
[12] Yasuhara K., Hirao K. and Hyde A.F.L. (1992).
Effects of cyclic loading on undrained strengthand compressibility of clay. Soils andFoundations 32 (1), 100-116.
[13] Moses G. G. and Rao N. S. (2003). Degradationin Cemented Marine Clay Subjected to Cyclic
Compressive Loading. Marine Georesourcesand Geotechnology. 21(1), 37-62.
[14] Wang S., (2011). Postcyclic behavior of lowplasticity silt. Department: Civil, Architecturaland Environmental Engineering.
Doctoral
Dissertations.Paper 2009. Missouri Universityof Science and Technology

Table 2 Initial stiffness and shear strength results
for static and post-cyclic

Test
StaticPostcyclic
Reduction(%)

Location

Initialstiffness (Ei)Shearstrength (Sy)Lumadan

26.66

29.37

Parit Nipah

15.76

46.02

Lumadan

10.45

21.55

Parit Nipah

10.73

18.35

Lumadan

60.80

26.63

Parit Nipah

31.92

60.13

Yield shear strength decreases by the number ofcyclic loadings, and exceeds a failure limit in 100
cycles with 9% and 12% of an axial strain for
Lumadan and Parit Nipah peat. Allowing the deviator
stress developed an increase during post-cyclicloading, initial stiffness decrease significantly andmore pronounced in knee point at the axial strain of
only 1.4% and 1.5% intersect in turn the slope of thecurve of deviator stress until the end of the test.
CONCLUSION
In summary, the main contributions to research in
the determination of the post-cyclic yield shear
strength and initial stiffness of Lumadan and Parit
Nipah peat soil can be included as follows:
1. The post cyclic yield shear strength ofundisturbed peat soil is considerably lower
compared to static while cyclic loading are
imposed to specimens after 100 numbers ofcycles.
2. The initial stiffness of post cyclic of peat soil
is governed by the softening behaviour that
occurred when the specimen exhibited ±1.5% axial strain.
3. The level of post-cyclic yield strain occurredin ±0.62% of axial strain for Parit Nipah peatand ±0.58% for the undisturbed peat soils.
4. Cyclic loading causes deformation of soil
structure and alteration of void will resulted
in decreasing of yield shear strength.
REFERENCES[1] Huat, B. B.K. (2004). Organic and Peat Soils
Engineering. Universiti Putra Malaysia Press,
Serdang, Malaysia, pp. 20-80.
[2] Wong L.S, Hashim R. and Ali F.H (2008).
Strength and Permeability of Stabilized PeatSoil.
[3] Deboucha S. Hashim R. and Alwi A. (2008).
Engineering properties of stabilized tropicalpeat soils. Electron. J. Geotechn. Eng., 13E.
329

GEOMATE- Bangkok, Nov. 14-16, 2016[18] Zainorabidin, A. and Bakar I. (2003).
Engineering Properties of in-situ and Modified
Hemic Peat Soil in Western Johore. In
Proceedings of 2nd International Conference on
Advances in Soft Soil Engineering andTechnology (pp. 252 - 261). Putra Jaya,
Malaysia.
[19] Michael T. H., Jitendra S. S. C. Derek M. S. and
Lee B. (2011) Effect of fibre content andstructure on anisotropic elastic stiffness andshear strength of peat. Canadian Geotechnical
Journal, 2012, 49(4): 403-415, 10.1139/t2012003.

[15] Wood, D. M. (1990). Soil Behavior and Critical
State Soil Mechanics, Cambridge University
Press.
[16] Adnan Z. and Habib M. M. (2015). Pre- andPost-Cyclic Behavior on Monotonic Shear
Strength of Penor Peat. Electronic Journal of
Geotechnical Engineering. Vol. 20 (2015),
Bund. 15. 6928.
[17] Zolkefle S. N. A. (2014). The Dynamic
Characteristic of Southwest Johor Peat underDifferent frequencies. Degree of Master in Civil
Engineering Thesis.

330

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

PROMOTION OF ICT UTILIZATION BY ELECTRIC RESISTIVITYMANAGEMENT IN FLUIDIZATION TREATMENT PROCESS FORGROUND IMPROVEMENT
Yasuhide Mochida 1, Indra Hardi 2
Professor of Department of Architecture and Urban Design, Graduate School of Science and Engineering,
Ritsumeikan University, Japan; 2Master Degree Student of Department of Architecture and Urban Design,
Graduate School of Science and Engineering, Ritsumeikan University, Japan

1

ABSTRACTThis research introduced a new method for monitoring the quality of ground improvement body by measuring
the electrical resistibility of the ground improvement body using the ICT device attached on the bucket mixer. Thismethod shows the engineers the unevenness of electrical resistivity of the ground improvement body, which means
imperfection in mixing process or inconsistency of the compound within the ground improvement body. The
monitoring is intended to perform real-time two-dimensional cross-section observation of the ground improvement
body. This method is expected to be useful to minimize the vague judgment in the decision-making process on the
ground improvement stirring states.
Keywords: Ground improvement, electrical resistivity, shallow foundation, ICT
INTRODUCTION
In application of the fluidization treatment
method for the ground improvement work, the mostcommon method to make the final decision to ensure
the performance of the ground improvement body areby confirmation of the rotation number or mixing
device’s tachometer and flow value of the stirrer
control and also by visual observation of theengineers on the color and consistency of the mixture.
However, it is difficult to confirm the entire parts ofground improvement body and it cannot be denied
that there is a possibility of inhomogeneous mixture
inside the ground improvement body.
With the utilization of the ICT (Information andCommunication Technology), by performing
management procedure by attaching an electricalresistivity sensor on the bucket mixer and measure theelectrical resistivity of the ground improvement body,
it is possible to monitor and determine the unevennessof the ground improvement body mixture andunderstand the performance of the ground
improvement body in real-time.
FLUIDIZATION TREATMENT METHOD OFGROUND IMPROVEMENT

determine the design for the compound to be used on
ground improvement work. Then perform
construction plans at the execution stage, such as
construction order and quality control and continued
by inspection stage and completion of theconstruction process.

Fig. 1

Ground improvement work flow

Figure 2 shows the flow chart of the execution
stage, which are describes as follows.
1.
2.

Confirmation of ground improvement positionby planting steel rod in the boundary area..

Removing excessing unsuitable for construction
soil by excavation work.
3. Excavation of soil within the boundary area of
the ground improvement site to the expected
depth.
4. Verify the support soil/ground by measure thegeometry of the cavity. Determine thefluidization treatment agent volume to be use
based on the calculated cavity’s volume

Figure 1 shows the flowchart of the ground
improvement work. The first step is to perform afeasibility study such as building's shape, necessary
loads to be transferred directly to the foundation andsoil properties to perform a design process for ground
improvement. Subsequently, perform an indoor
compound examination using local materials and331

GEOMATE- Bangkok, Nov. 14-16, 2016

5.

Decide the reference line on the ground
improvement area range using a backhoe and thebackhoe’s tip as the reference

described by equations (1). With the measurement
units follows; voltage (V): V, resistant (Ω): R, andelectric current (A): I.
Voltage ∶ E V

R

IElectric current ∶ I A
Resistance ∶ R Ω
E/I(1)
Figure 3 shows the cone type sensor (4-electrode)that was used in a mixing investigation. This cone
type sensor flows the electric current (A) from two of
the four electrodes (outside electrodes) while the 2electrodes located inside measuring the voltage. Withknowing value of current and voltage, electric
resistant and electrical resistivity can be calculated.

Fig. 3 Cone type sensor (4 electrodes type)Fig. 2 Execution stage
6. Fluidization treatment process (construction)
7. Confirm the improved level of the ground
improvement body.
8. Electrical resistivity measurement.
9. Sample collection for laboratory test.
10. Completion of the process.
ELECTRICAL RESISTIVITY APPLICATIONThe Degree of Mixing Process Investigation by
Electrical Resistivity Measurement
This investigation method is using themeasurement of electrical resistivity of the soil andsolidifying agent compound and estimates the mixing
degree from the change of the electrical resistivity ofthe parts of ground improvement body.
Electrical resistivity is the opposite of the electric
current flow, which is expressing the resistor forelectric current. Cement, waterglass injection,
bentonite clay, etc., are the main material used forconstruct the ground improvement. If the ion
concentration of those materials is high, the electricalresistivity became lower and changes because of themixing process with the soil. Therefore, if thecompound perfectly stirred/mixed the electricalresistivity become lower and when there are a highvalue of electrical resistivity it may indicate that there
are slumps or incompletely stirred compound at thatpart. This method allows engineers to verify themixing degree of the ground improvement body withless time needed.

The cone type sensor have a diameter
approximately 25mm, with the length of the sensor
unit of 100mm, the electrode thickness of 5mm andspacing between the electrodes are 20mm.
Measuring Method (Insertion Rod Measurement
Method)Measurements of the electrical resistivity perform
immediately after completion of stirring process orafter the expected rotation counter number is satisfied,by sinking the cone sensor at certain points on thedepth direction. Investigation method performs bysuspending the sensor on the hydraulic shovel
available at the construction site, and penetrates at aconstant speed. Investigate two or three points in one
area and control by make relationship between
average moving value and depth distribution andbetween average electrical resistant value and depth
distribution. The interval of measurement points is set
to 10cm while one side of the investigation point
decides using 1meter length triangle ruler marker andthree investigation points decide using the range of
50cm equilateral triangle ruler.
Application Example of Construction
Management by the Electric Resistivity
Utilization Method
The examination conducted in five locations withdifferent type of soil; Kanto loam soil on two location(Chofuyagumodai-Tokyo, Tokorozawa-Saitama (Prefecture), sandy silt soil on one location (NagaiYamagata Prefecture), decomposed granite (DG) soilon one location (Koriyama-Fukushima Prefecture)and clay soil on one location (Nagaoka City-Niigata
Prefecture).

Electrical Resistivity Ratio
According to Ohm’s law electric resistant ρ
332

GEOMATE- Bangkok, Nov. 14-16, 2016

coefficient of variation for the compressive strength
was set at 30% as a target and experiment results has
satisfied the target. On the other hand, the trend ofcoefficient of variation of the electrical resistivity is
not due to the soil type, and the result satisfies thevalue of 20%.

Fig. 4 Vessel Test Result
Cement-based solidifying material has been used
to the ground improvement as an additive material
with the addition of sandy silt and Kanto loam soil
with an amount of 200~350 kg/m3. The basic process
for ground improvement on this research is byperformed solidifying process to several soil layersthat need to improve and finally performed thefluidization treatment process to the entire part ofground improvement body. Vessel test has been
performed at each position in order to set the target
value for the mean electrical resistivity value due to
fluidization process.
According to vessel test results (Figure 4), thetarget of electrical resistivity value of the improved
body was set on 10 ~ 20 (Ω · m) and the average
resistivity value of each investigation points has
satisfied the expected value. The examination resultsin DG soil on Fukushima Prefecture and Kanto loam
on Tokyo shows the big difference of average
electrical resistivity value partially. Thus, there wasimperfection of the stirring process and the valuechanged into expected range after another stirring
process.

Fig. 5 Relation between electrical resistivity value
and cement amount
Relationship between Compression Strength of
Ground Improvement Body and Electrical
Resistivity Value of Field Core.
Figure 6 shows relationship between strength of
the improvement of the electrical resistivity and thefield core. The strength of ground improvement bodyon full-length core is using uniaxial compressive
strength after curing (28 days) and electricalresistivity of the location where the sample foruniaxial compressive was taken is used for electricalresistivity value. If the linear approximation formulaand regression analysis in each three types of soil iscalculated, the following formula is derived.
989.1
223
(2)
12.7
3272
(3)
42.3
3087
(4)
Compressive strength is increased as electricalresistivity value increases is the mutual trend of every
type of soil. In particular, this tendency is significant
when sandy soil (Equation 2) is compared with clay
and loam soil.

Variation of the Electrical Resistivity of the Final
Stage in the Ground Improvement Work
Figure 5 shows the relationship between strengthof ground improvement body and the coefficient ofvariation of the electrical resistivity value of the sitecore. Local soils are selected for experiment fromtotal 31 locations: sandy soils (sand from threelocations, DG soils from one location, fine sands fromone location), clay group (clay from 11 locations,
sandy silt from six locations, silt from three locations)and loam (Kanto loam from six locations). The
strength of ground improvement body mixture in thesite core was calculated using coefficient of variationof unconfined compressive strength after curing(material age of 28 days). Coefficient of variation ofelectrical resistivity is defined as coefficient ofvariation based on electrical resistivity value where
the sample for compressive strength test was taken. In
this construction method of ground improvement,

Fig. 6 Relation between electrical resistivity value
and compressive strength of ground improvement
body
ELECTRICAL RESISTIVITY APPLICATIONOF ICT UTILIZATION IN CONSTRUCTION
MANAGEMENTElectrical resistivity measurements in the past
research gave an example how to understand theground properties by gathering electrical resistivity
333

GEOMATE- Bangkok, Nov. 14-16, 2016

value at a specific location using the stationed
electrical resistivity measurement sensor. The
technology used in this research provides a method in
how to understand the ground properties on a twodimensional cross section of ground improvement
body by measuring the compound’s electricalresistivity using electrical resistivity sensor attachedon moving bucket mixer in order to perform a stirring
process. This paper shows the basic concept ofpromoting ICT technology described as follows.
• Made the possibility to observe the visuallyinvisible underground situation• Speeding up the performance identification by
utilize the electric signal•Allows instant observation by display method
Therefore, this technology application capable ofenhancing certainty of ground improvement
performance identification by eliminate the lack of
the information on the performance on theconventional method by human eye visualobservation and improve productivity by eliminating
the ambiguous judgment of the stirring rotation
counter number method. Finally, this technology is
expected to reduce the misjudgment on the
management process made by recent lack of skill
engineers.

Fig. 8 Bucket mixer configuration
Calibration
Calibration procedure for the method andequipment describes as follows.
1. Put distilled water in the mud balance instruments
to manage the specific gravity of the solidifying
material liquid, ensure the specific gravity value is1.0. Measure the specific gravity of the solidifying
material liquid measured in mud balance
instruments and ensure the specific value is 99%.
Ensure the pail if the bucket mixer in thepredetermined shape.
2. Store the solidifying material liquid into the vessel
and ensure the value of electrical resistivity is lessthan 10Ω.
3. Inspection of the horizontal position movement andorientation of the bucket mixer by actual
measurement and monitor observation as areference and ensure the different is ± 5cm in 2
meters horizontal movement. As well for
inspection of vertical movement, the different is ±5cm in 1meter vertical movement.

Fluidization Treatment Process of Ground
Improvement Construction Management
Figure 7 shows diagram of wirelesscommunication system and Figure 8 shows diagramof attached electrical resistivity sensor on the pail partof the bucket mixer. Mixing rotor and solidifying
material liquid discharge device configures the pailpart of the bucket mixer device to be used in thisconstruction method. Furthermore, the pail part isequipped with depth gauge, tachometer, and anelectric resistivity sensor, which has a function fortransmitting the mixing situation of the mixture inside
the ground improvement body to the surface.

Monitoring
Data including compound uniformity level
displays on the monitor in real-time to be use for
evaluating the process. Monitoring situation andelectrical resistivity distribution show in photo 1. The
electrical resistivity distribution in the entire ofground improvement body displays in the different
color according to its value. The ground improvement
body displays on cross section view in 25cm x 25cm
grid form and the measured electrical resistivity
appears in colored-dot shape. The stirring processcompleted if electrical resistivity distribution
displayed uniformly in pink or yellow dot indicators.

Fig. 7 Wireless communication system diagram
Monitoring procedure will be conduct by theoperator using electric signals sent from equipment
attached on the pail to the control unit. Operator
judges the compound mixture condition of the ground
improvement body based on information displayedon the monitor and able to understands the shape andthe compound uniformity degree of the ground
improvement body. Measured electrical resistivity
value will be display on the monitor screen on
different colors based on its value and divide theground improvement body into 25cm x 25cm grid

Photo 1 Monitoring and the electric resistivity value
distribution

shape.
334

GEOMATE- Bangkok, Nov. 14-16, 2016

the pail part of the bucket mixer and performs
monitoring process.
3. Monitor observation.
Electric signals sent by the sensor transmitted to thecontrol unit cabin monitored. The operator will
understand the stirring state of the ground
improvement body compound by judging its
condition based on the information displayed on themonitor.
Electrical resistivity value varies depending on
the soil type, but in the homogeneous stirring/mixing
state converge within 7 ~ 20Ωm. Hence, maximum
value is set on 20Ωm. If the measurement value
exceeds 20Ωm or there is significant variation seen in
the resistivity value of the measurement in the depth
direction, the existence of not completely stirred partis become a concern. In this occasion, re-stirring
process must be performed immediately until theelectrical value uniformly distributed within the
expected range. Subsequently, perform a quality
control process by measuring electrical resistivity
value in the depth direction by inserting electricalresistivity sensor using insertion rod into the ground
improvement body.

Control Device
Figure 9 and Figure 10 show the depth gauge used
for measuring the coordinates of the pail part ofbucket mixer and ground improvement traces
monitoring. The coordinate management proceduredescribes as follows.
1. Set the position of the pail from the length andangle of each arm using inclination sensor.
2. Define a reference line using one side of a backhoe
inside area range of ground improvement and set
the tip of the bucket mixer pail in this position.
3. From this position, calculate the pail part position
by angle and length of each arms and display the
trajectory.
4. Input the dimension of ground improvement area
range and depth calculated from reference line, anddisplay them on the monitor.
Ground improvement trajectory shows the outer
periphery of the body and it is assumed as the
maximum trajectory. Ensure that this maximum
trajectory sufficiently satisfied the ground
improvement body geometry by observing themonitor. Quality control performs by measuring
electrical resistivity inside the ground improvement
body using electrical resistivity sensor attached on the
tip of bucket mixer pail. Ground improvement crosssection displays into 25cm x 25cm grids pattern andthe electrical resistivity will be measured
continuously at each grid so the change of its value
can be observed in real-time. The construction control
procedure describes as follows.

QUALITY CONTROL OF UNSOLIDIFIEDSPECIMEN
Quality investigation of the ground improvement
mixture due to fluidization treatment process iscommonly conducted by performing unconfinedcompressive strength test of the specimens taken fromsolidified ground improvement body using core cutter
or core boring. This method allows engineers tounderstand the performance of the ground
improvement body at early stage and assuming thenext progression.
However, in this research, sample is collected bycollecting unsolidified specimens from ground
improvement body and also allows collecting
specimens by using both of remixing mold core andinsert mold core. Subsequently perform comparisonof compressive strength between remixing core andinsert core, and estimating the strength of boring core
due to remixing core, and shows that proposed
method collecting sample using mold core wasreliable.

Fig. 9 Coordinated measurement

Experiment on Performance Confirmation of
Unsolidified Sampler

Fig. 10 Ground improvement trajectory
1. Electrical resistance measurement of solidifying
agent liquid.
The value has tendency express the compoundstirring state. Measure the solidifying agent resistivity
prior to examination and assume it as the minimum
value.
2. Electrical resistivity measurement of ground
improvement body.
Measuring the electrical resistivity value of the entire
ground improvement body, using a sensor attached onExperiment on performance confirmation isconducted in total 9 locations, which are loam in 6
locations, sandy silt in 1 location, silt in 1 location andsandy clay in 1 location. Terms and conditions ofeach of the ground improvement are W/C = 80 ~ 150(%), additive amount of 250 ~ 300 (kg/m3), design
strength criteria of 450 ~ 1050 (kN/m2) and thethickness of ground improvement body from 1.6 m to335

GEOMATE- Bangkok, Nov. 14-16, 2016

4.5 m. As a reference, 3 locations from total 9
locations of boring core and remixing core was
performed boring core (48φ ~ 72φ), remixing core(50φ) and insert core (100φ). Both of boring coreand mold core was subjected to compressive strength
test in moist air (20 ℃ ± 3 ℃) based on Japan
compressive test standard (JIS A 108).

CONCLUSION
This research showed an example of practicalutilization of ICT technology in construction
management by investigating the two-dimensionalcross-section monitoring to observe the mixing
degree of the entire ground improvement body inreal-time by electrical resistivity value measurement
using devices attached on the pail part of bucket
mixer. The technology is aiming for the realization of
production improvement with adding a user-friendly
device to deal with ambiguous judgment or inaccurate
observation result made by the engineers in verifying
the ground improvement body stirring states.
This research has proved that quality control using
electrical resistivity measurement of ground
improvement body provides great accuracy due to 2
dimensional monitoring of ground improvement
body along with dot colored indicators that indicateselectrical resistivity value of ground improvement
body mixture.
Furthermore, by understanding the differences inelectrical resistivity value due to soil type,relationship between cement amount and electricalresistivity value, and relationship between strength ofground improvement body and coefficient ofvariation of electrical resistivity, this method has
satisfied targeted value of 30% on both of coefficientof variation and quality control.
In addition, due to mold core on unsolidified
sample of ground improvement body and performed
compressive strength test at early stage, with thesample taken at any depth, allows engineers toestimate the strength of ground improvement body at
any material age.

Strength comparison of boring core and the moldcore

Fig. 11 Comparison of average compressive strength
between mold core and boring core(remixing mold core)
Figure 11 shows the comparison of averagecompressive strength between mold core (remixingcore) and boring core. The average compressive
strength of mold core is roughly considered about
80% of the average compressive strength of theboring core.
Strength Comparison of Remixing Core andInsert Core
Figure 12 shows the comparison of averagecompressive strength between mold (remixing, insert
core) and boring core. As been shown in previous
research on compressive strength of the concrete core,
the compressive strength and coefficient of variationof core’s strength increase due decreasing in core’s
diameter, and on concrete core with maximum
aggregate size (5 mm to 20 mm), its compressive
strength decrease due to increasing in core’s diameter.
Compressive strength test result in this research also
shows the trends that compressive strength of theinsert core has more variation and compressive
strength of remixing core has the higher value.

REFERENCES[1] “Design ground improvement and quality
control guidelines for buildings”, Japan
Architecture Center,2006[2] Sakai Y, “Quality control method (undergroundstake and consecutive underground walls) using
electrical resistivity”, Japanese GeotechnicalSociety, July 2007, the 42nd conference[3] Mochida Y, “Quality control method (ground
improvement and fluidity processing) using
electrical resistivity”, Japanese GeotechnicalSociety, July 2007, the 42nd conference[4] Mochida Y, “A study about electric ratio
resistance and it relations to mixture degree”,
Japanese Geotechnical Society, July 2008, the43rd conference[5] Tamura T, “Field experiment on the quality
evaluation of improved body by electricalresistivity”, Architectural Institute of Japan,
May 2000, structure system papers no.531

Fig. 12 Comparison of average compressive strength
between mold core (remixing, insert core) andboring core336

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

GROUT PROPERTIES OF MICROPILE UNDER DIFFERENT SOIL
CONDITIONS
Putli Yasmeen Nadeah M. Y.1, Hassanel Zachary A.2 Julinda Hena J.3 Siti Aishah M. S.2
1

Mariwangsa Maju Sdn. Bhd., Malaysia ; 2Faculty of Engineering, Universiti Malaysia Sabah, Malaysia ;3
STL Geotechnical Engineering Sdn. Bhd., Malaysia.

ABSTRACT: Micropiles application have grown widely as a foundation options especially for underpinningof existing structures, strengthening of foundations and supports for buildings and bridges. Buried
underground, the grout of micropiles may encounter aggressive environment that may detrimental to its
properties. The essential of this research is due to the increasing number of micropiling applications throughout
countries and to investigate the grout properties of micropiles under different soil conditions through soil
sample testing, water chemical testing and grout cube testing. This paper found that grout specimens cured intreated water which have higher sulfate content have the highest percentage of ettringite with 43.18% and 20%of the percentage of hydration products, specifically portlandite (Calcium Hydroxide) are the lowest comparedto the other curing conditions. The formation of ettringite may develop tensile stress and micro cracks inside
the grout cube, thus lead to strength loss and allow more sulfate ingress that can add into further deterioration.
Conclusively, the grout cubes cured in untreated water with lower sulfate content are more likely to produce
higher early and 28 days compressive strength.
Keywords: Micropile; Foundations; Grout; Sulfate;
1. INTRODUCTION

2. METHODOLOGY

Micropiles had been used throughout the world
since their development in Italy in 1952 (FHWA,
1997). The micropiling method was introduced in
Malaysia in the early 1980s and gained
acceptability since then particularly in difficult
ground conditions, limited working area andtreacherous limestone areas. However, there is only
limited number of studies concerning micropiles
especially in Malaysia.

In order to imitate the actual environments thatwill surround the micropiles grout, 16 test cubes areprepared and cured into 4 different conditions. The
conditions are as follows:i.
Condition 1: Curing of 4 nos of test cube;
water sourced from the daily water supply (treatedwater)
ii. Condition 2: Curing of 4 nos of test cube;
water sourced from river (untreated water)iii. Condition 3: Curing of 4 nos of test cube;
sandy soil and immersed with water sourced fromriver (untreated water)iv.
Condition 4: Curing of 4 nos of test cube;
sandy silt soil and immersed with water sourced
from river (untreated water)

The micropile must be designed to satisfy thegeneral criteria of the foundations design which it
must be located properly so as not to be adversely
affected by outside influences (Liu and Evett,2005). The outside influences that may affect thefoundations are water, significant volume change,
underground defects (caves, for example), frost andgroundwater, adjacent structures and property lines.
In accordance to the micropile design, grout plays
the critical part of micropiles during its lifetime.
Nonetheless, the study on micropiles and laboratory
data are very few, thus the behaviour of the grout
properties of micropiles under different soil
conditions remains unclear.
The aims of this research are to study themechanical properties of micropiles grout subjectedunder different soil conditions. After being casted
in different soil conditions, the microstructures ofthe micropiles grout samples will be analyzed to
study the effects of different environment towards
its composition.

Fig. 1: (Top Left) Curing Condition 1. (Top Right)
Curing Condition 2. (Bottom Left) Curing
337

GEOMATE- Bangkok, Nov. 14-16, 2016

Condition 3. (Bottom Right) Curing Condition 4.

RESULTS AND DISCUSSION

Laboratory test conducted are divided into threegroups namely soil sample testing, water chemical
testing and cement grout testing. The grout cube
testing is performed on 16 cement grout cubes bothfor 7 days and 28 days curing period. For soil
sample testing, two soil samples were taken foranalysis. There are four major tests conducted forthe soil samples such as Sieve Analysis,
Hydrometer Test, Atterberg Limits and Specific
Gravity.

3.1 Soil Sample Testing

For water chemical testing, both treated anduntreated water samples are analyzed in terms ofsulfate content and chloride content. Water
chemical testing were conducted to study the effectof the chemical contents in the two water samples
towards the grout cube properties. The waterchemical testing includes the Sulphate Content and
Chloride Content.

3.1.1 Sieve Analysis
Based on visual observation, the soil samples forthe purpose of conditions preparations as mentioned
in methodology of research are expected to be sandy
soil and clay soil. The position and shape of thegrading curve determines the soil class and the soil
sample No. 1 are classified as gravely sand.
3.1.2 Hydrometer Test
This soil sample No.2 are considered to be clay
based on visual observation. The wet sieving
procedure separate the silt and clay sized particles
for hydrometer testing. The value of percentage
finer (passing) is then plotted in the particle size
distribution graph. From the graph, it was found thatthe soil samples are categorized as sandy silt. The
soil sample can be further categorized by texturalclassification by referring to USDA Soil Textural
Triangle, thus can be classified into silt loam
category.

Finally, the parameters of grout cubes are
obtained from bulk density test, compressive
strength SEM images and X-Ray Diffraction
analysis for all grout cubes.

3.1.3 Atterberg Limits

Table 1 Grout cube tagging

Table 2 Summary of Attenberg limit test data

Condition

Days of
Curing1(Treatedwater)

7

2(Untreatedwater)
3(Untreatedwater +
sandy soil)
4(Untreatedwater +
sandy silt
soil)

28
7
28
7
28
7
28

Cube Tag
C1a
C1b
C1c
C1d
C2a
C2b
C2c
C2d
C3a
C3b
C3c
C3d
C4a
C4b
C4c
C4d

Summary
Liquid Limit
Plastic Limit
Plasticity Index

35
14
21

The Atterberg Limits are carried out by
Cassagrande method. The soil sample used for thecondition 4 (Sample No. 2) has the average of
Plastic Limit of 13.55% whereas the Liquid Limit is34.16%. The soil samples are categorized as organic
silt based on USCS (ASTM D 2487).
3.1.4 Specific Gravity
The first soil sample, Sample No. 1 iscategorized as Sandy Soil and the value of thespecific gravity is assumed from the range of the
sandy soil which is from 2.63–2.67. The range forthe silt soil type is 2.65-2.7. From this test, thespecific gravity for this sandy silt soil type is 2.655
which are under the range of silt soil specificgravity.

3.

338

GEOMATE- Bangkok, Nov. 14-16, 2016

3.2 Water Sample Testing

After 7 days of curing in the 4 conditions as
specifically prepared in this research, 2 cubes foreach condition are crushed to determine thecompressive strength. It was found that the grout
cube C3b which is cured in untreated water andsandy soil has the highest compressive strength of
69.98 MN/m2. On the other hand, the grout cube
which has the lowest strength is C1a with 53.95
MN/m2 which is cured in a conventional methodfrom daily water supply.

In this research, there are two types of water
sample used in the 4 curing conditions of the grout
cube which is from treated water and untreatedwater, which is sulfate and chloride content. The
chloride content of both water samples is nearly thesame which is 23 mg/L and 24 mg/L. whereas thesulfate content of the untreated water is lower
compared to treated water. The sulfate content in
the treated water is relatively higher, could be
resulted from the use of Aluminium Sulfate fromthe water treatment plant.
3.3 Grout Cube TestingCompressive Strength (MN/m2)

In this research, there are a total of 16 nos. of
micropile grout cube that will be analyzed
according to the bulk density, moisture absorption,
compressive strength, microstructure and chemicalcomposition.
3.3.1 Bulk Density
For the grout cube with 7 days of curing, it was
found that grout C3b has the highest bulk density
which is 1.981 Mg/m3 whereas grout C3a has thelowest value with 1.86 Mg/m3. Meanwhile for 28
days of curing, grout C4d has the highest bulk
density which is 1.989 Mg/m3 whereas grout C3a
has the lowest value with 1.877 Mg/m3.

Fig. 33.3.2 Moisture Absorption

Compressive Strength MN/m2

74.15
59.15

C1c C1d C2c C2d C3c C3d C4c C4d
Grout Cube
Graph of compressive strength of grout
cubes (28 days curing)

By referring to the grout cube averagecompressive strength, the cubes which is cured in
Condition 2, 3 and 4 is higher than cubes cured in
Condition 1. In general, the grout cubes cured in
untreated water with lower sulfate content are more
likely to produce higher early and 28 days’
compressive strength. The possible cause for lower
strength for grout cube cured in treated water may
be due to the formation of expansive compounds asa result of chloride and sulfate attack. Yet, the longterm performance of the grout cubes cured in thiswater samples are still uncertain.

3.3.3 Compressive Strength

Fig. 258.9458.23
55.3454.49
49.6

66.8

At 28 days of curing, it is found that grout cube
c4d which is cured in condition 4 (untreated waterand sandy silt soil) has the highest compressive
strength with 74.15 MN/m2. The cube with thelowest strength is C3c, 49.6 MN/m2 which is cured
in untreated water and gravely sand.

At 7 days curing ages, grout C3a has the lowest
moisture absorption with 1.45% while C4b has thehighest value with 3.03%. For the 28 days of curing,it showed C3d has the lowest moisture absorption
with 1.44% while grout cube C4c has the highest
value with 2.80%.

80
70
60
50
40
30
20
10
0

80
70
60
50
40
30
20
10
0

Compressive Strength(MN/m2)Compressive Strength(MN/m2)
69.98
69.65
61.61
58.08
56.13
53.95
52.49
42.81

C1a C1b C2a C2b C3a C3b C4a C4b

Grout Cube
Graph of compressive strength of grout
cubes (7 days curing)339

GEOMATE- Bangkok, Nov. 14-16, 2016

3.3.4 Microstructure (SEM & EDX)

From SEM imaging of the grout cube
specimens, the ettringite formation is present inalmost all grout cube specimens. Nonetheless, theintensity of the entringite formation may varydepending on the curing conditions. Additionally,
the results and severity of the sulfate attack might
be different with higher sulfate concentration.

Fig. 4 The overview of the microstructure ofgrout sample from grout cube C1a and C1b
From the overview, the grout of C1b has arelatively smaller void than grout C1a. Meanwhile,the distant view of specimens C1d shows a great
deal of larger air void.
Fig. 6 Formation of ettringite inside an air voidin specimen C2d
3.3.5 X-Ray Diffraction and Quantitative Phase
Composition of Grout Specimens
For the phase identification to determine thechemical compositions of the grout cube specimens,
each X-Ray patterns of the 16 grout cube specimens
are analyzed by using software and the phaseassumed to be present will be selected based on the
figure-of-merit and visual inspection of theiragreement with the experimental pattern.

Fig. 5 Ettringite formation in specimen C1a
Generally, in cement-based material, sulfate
attack mainly occurs from the exterior andinternally. External attack arises from increased
sulfate sulfate concentration from the outsideenvironment (Menendez et. Al, 2013) and is morecommon and occurs when water contain dissolved
sulfate penetrate or come into contact with thecement-based material. The microstructure of thesample experiences some changes that vary and thecommon signs are the formation of ettringite.

In this analysis, it is important to pay attention
to the hydration products and the percentage of the
expansive crystalline presents in the specimen as
both of these are the main factors in the mechanismsof the external sulfate attack. In external sulfate
attack, the waterborne sulfates from medium suchas soil and water, reacts with hydration products toform expansive crystalline products such as the
ettringite, monosulfate and gypsum.

In this research, ettringite formation is observed
from SEM and confirmed by EDX. From larger
SEM magnification, it was found that from sample
C1a, show presence of prismatic fine needle shaped
crystals as shown in figure below. The structure is
identical to the ettringite that are commonly foundin cement based materials.

Table 3 The average percentage of phase present ingrout cube specimens

SampleEttringite(%)

Gypsum(%)

Portlandite(%)

Condition 1
Condition 2
Condition 3
Condition 4

43.18
33.43
30.43
28.30

17.13
14.57
9.93
16.90

20.20
29.33
34.25
35.00
340

Tricalcium
Aluminate
(%)
3.33
4.67
1.00
3.38

Thaumasite(%)

Calcite
(%)

1.28
1.60
1.08
1.25

14.88
16.40
23.33
15.20

GEOMATE- Bangkok, Nov. 14-16, 2016

From the table, the average of each phaseobtained from 16 grout cube specimens andcompared with respect to the curing conditions. It isfound that grout specimens cured in Condition 1
possess the highest percentage of ettringite with
43.18%. Correspondingly, the percentage ofhydration products, specifically portlandite(Calcium Hydroxide) are the lowest with 20.20%
compared to the other curing conditions. Thisshows that most of the hydration products havereacted with waterborne sulfates resulting in thehigh percentages of the ettringite and low
percentages of portlandite.

a)Primarily, the micropile grout is a setting

material, normally cement and water, sometimes
containing additives or a limited amount of fineaggregates that transfers load from the bearing
element. During its service life, deterioration may
occur as a result from aggressive chemical
penetration into the grout. It is important to assess
the type of exposure, strength and microstructuralin order to achieve the specified micropile grout
criteria.
b)

There are many factors that may cause

deterioration and degradation of the micropile
grout, therefore, consideration should be taken toagents such as water and soils as it may containaggressive chemical that can reacts with theunstable cement phases. The performance of thegrout mainly depends on the environmental
exposure, chemicals that may diffuse through its
mass and the design quality.

Comparatively, grout specimens cured in
Condition 4 have the lowest percentage of ettringite
with 28.30%. Correspondingly, the percentage ofhydration products, specifically portlandite(Calcium Hydroxide) are the highest with 35%
compared to the other curing conditions. Thisshows that most of the hydration products have yet
to react with waterborne sulfates resulting in thehigh percentages of the portlandite and low
percentages of ettringite.

c) In terms of compressive strength, thecubes which is cured in Condition 2, 3 and 4 ishigher than cubes cured in Condition 1. Therefore,
the grout cubes cured in untreated water with lower
sulfate content are more likely to produce higher
early and 28 days compressive strength.
d)Based on the ACI 318 exposure classes,

the sulfate content and the chloride content in thewater used as curing conditions are negligible. The
sulfate exposure in this research is less than 150
ppm and the chloride content is less than 0.30 % as
specified by ACI. Therefore, the sulfate andchloride exposure in all the curing conditions gave
not as much effect to the grout cube properties.

Fig. 7 Examples of Micro Cracks as found inSpecimen C1a (Left) and C4a (Right)
Apparently, the average of the compressive
strength for grout cubes cured in Condition 1,
Condition 2, Condition 3 and Condition 4 are 55.47,
57.82, 60.63 and 61.44 MN/m2 respectively. When
comparing the phase identification results to thecompressive strength, it shows that the percentagesof ettringite presents in the grout cube relatively
affecting the performance of grout cube in terms ofstrength. The formation of ettringite may develop
tensile stress and micro cracks inside the grout cube,
thus lead to strength loss and allow more sulfate
ingress that can add into further deterioration.

5. ACKNOWLEDGEMENTS
Above all, I would like to offer my sincerest
gratitude to my supervisor and co author, Mr.
Hassanel Zachary Amaludin for his kind help,
motivation and effort. Very special thanks to STL
Geotechnics for the generous cooperation andcollaboration in this research and Engineer Miss
Julinda for taking time from her busy schedules toadvise me on my project. The technical supports
and encouragements offered are very crucial
throughout the completion of this research. Not toforget, to both of my parents and family members
for the unconditional support and understanding.

4. CONCLUSION
In this research, consideration has been given to
the grout properties of micropile in different soil
conditions. The following are the general
concluding remarks to define this research:
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GEOMATE- Bangkok, Nov. 14-16, 2016

6. REFERENCES(2013), Sulfate Attack of Concrete,[1] FHWA Micropile Design and Construction

Performance of Cement-Based Materials in
Aggressive Aqueous Environments, 10(1), 774.

Guidelines Implementation Manual Priority
Technologies Program (PTP) Project. 2000.
Washington: Publication FHWA-SA-97-070.

[4] ACI 318: Building Code Requirements for[2] C. Liu and J. Evett. 2005. Soils andReinforced Concrete. 2008. American
Concrete Institute, Farmington Hills.

Foundations SI Ed.: Prentice-Hall.

[3] E. Menendez, T. Matschei and F. P. Glasser

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EXPERIMENTALLY MEASUREMENT AND ANALYSIS OF STRESSUNDER FOUNDATION SLAB
1

M. Mohyla1 and K. Vojtasik1 and M. Stolarik1 and M. Pinka1 and H. Lahuta1
Faculty of Civil Engineering, VSB - Technical University of Ostrava, Ostrava, Czech Republic

ABSTRACT
Understanding of a load redistribution into subsoil below building foundation is an important knowledge forreliable design and its economy too. The article presents the results of a physical model of a foundation slab andits interaction with the subsoil. The interactions were investigated comprehensively by monitoring thedevelopments of stress in the subsoil and foundation slab settlement during its loading. The load acting on thefoundation was applied by strutting the hydraulic press against heavy steel frame which was established by the
Department of Building Structures, Faculty of Civil Engineering of VSB -TU Ostrava for this purpose. The
preparatory phase of the present experiment involved the homogenization of soil during which trio pressure cells
in three horizons were gradually fitted. The quality of homogenization was checked on an ongoing basis throughfield tests: dynamic penetration load test, dynamic plate load test and seismic measurement of foundation slab
response. Finally, the homogenized soil was subjected to mechanical analysis to determine the strength anddeformation parameters for basic Mohr-Coulomb constitutive model.
Keywords: Physical model, foundation slab, foundation-subsoil interaction, subsoil pressure

may be installed gradually in the course of subsoil
creation. This paper deals with the direct
experimental measurement of stress below thefoundation slab, as this measurement was performed
on a real foundation built on artificial subsoil. The
paper documents in detail the preparation andverification of the quality of artificial subsoil,a system of measuring pressure cells installed atthree horizons, conditions during measurement, therecorded data, and data processing.

INTRODUCTION
Surface foundation slab transfers the load fromthe building into the subsoil (or foundation
environment). The load is transferred into thesubsoil by contact stress on the physical surface
interface between the foundation slab and the subsoil.
From this interface, stress extends further into thesubsoil and the stress increment in the subsoil
caused by the load is progressively reduced withincreasing distance from the physical interface. The
extent and value of stress increments below thefoundation in the subsoil caused by the load aredetermined by the appropriate analytical methods offorce balance, and deformation or numericalmethods. Direct stress measurements are only
carried out on the horizon of the foundation base in
the case of major building structures [1]-[3].
Experimental measurements on the horizons below
the foundation base were conducted rarely based onin situ physical models at 1:1 scale on the original
subsoil [4], [5]. So far, the calculation of stressvalues at different subsoil horizons below thefoundation has not been verified at points such as thecentre of gravity, characteristic point, edge andcorner of the foundation slab surface. The
installation of measuring pressure cells into theoriginal subsoil above each other at different
horizons is impossible without disturbing the subsoil
structure, thereby affecting measurement results.
Installing pressure cells distributed across the
subsoil in the desired scheme is only possible if the
subsoil is artificially created, where pressure cells

Fig. 1 Heavy steel test frame
EQUIPMENTThe Faculty of Civil Engineering, VSB-TU
Ostrava has an external heavy steel test frame used
to conduct experiments on building structuralelements. Figure 1 shows a photograph of the test
frame. The steel frame was designed by the343

GEOMATE- Bangkok, Nov. 14-16, 2016

Department of Building structures in cooperation
with Department of Geotechnics and UndergroundEngineering. The aboveground part of the test frame
is a steel frame assembled from steel I beams.
The underground part consists of two foundationwall strips which are, additionally, anchored by 5mlong steel tube micro-piles (89/10mm). Each of thewall strips is anchored into the subsoil by five
micro-piles. Three of them are vertical and the tworemaining ones are inclined at both ends of the wall
strip.
The total bearing capacity of the micro-piles intensile load is 1MN. A detailed diagram of the test
frame structure is shown in fig. 2.

test frame, which was defined by the position of theframe test foundation and the depth of the test frame
foundation base (fig. 1).
Backfill: excavated soil was gradually (inlayers with a thickness of up to 15cm) depositedback into space under the test frame. The backfill
was performed manually, with the stochastic
selection of sampling points from the excavated soil.
It was then evenly spread over the entire area under
the test frame. The thickness of layers was modified
as necessary depending on the desired location
horizon of pressure cells.
Compaction: compaction was performed
using vibration rammer with compaction force of
12.0kN (fig. 3). The quality of compacting of eachlayer was continuously verified by a light dynamic
plate (impact modulus of elasticity E vd [MPa]). In
the case of surface areas with lower parameters, thearea was re-compacted and verified.

Fig. 2 Structure scheme of the heavy steel test frame
PREPARATION OF ARTIFICIAL SUBSOIL
Fig. 3 Completion of sub-layers by compaction

For the progress of the experiment to be
successful, it was desirable that the subsoil
influenced by interactions with the experimental
model foundation slab showed homogeneous
properties (descriptive, stress and strain properties).
Partial results of experiments for example [6], [7]and subsequent field tests, however, showed that thesubsoil does not show such homogenous properties.
The inhomogeneity of subsoil properties was due tothe geological profile in the region of interest [8] andthe technology used to build the test frame
foundations. These foundations are monolithic andhave been built in a sloped excavation by depositing
concrete into the formwork. Backfill was built using
the original soil; however, it has not been compacted
to the original state, which had the largest impact on
the inhomogeneity of the environment. Figure 3
shows the heterogeneity of the original soil; the wallof the excavation contains the original subsoil andfills, and the side sections next to the concrete
foundations contain backfill of the sloped excavationfrom the time the test frame foundations were built.
Based
on
the
identified
circumstances,homogenization was performed as follows:
Complete excavation of soil from under the

As a result of the above manner to perform thebackfill, it was possible to build a foundation
environment with quasihomogeneous layers
exhibiting homogeneous properties, especially in thehorizontal plane. This led to the creation of anenvironment with the vertical direction of anisotropy.
To ensure a homogeneous isotropic subsoil, theexcavated soil (approx. 20m3) would have to behomogenized as a one part. Given that complete
homogenization would be time consuming and dueto the absence of the required technology (machines),
homogenization of the foundation subsoil was
performed as described above. The layout of theexperiment and the method of applying load on theexperimental foundation allowed the above methodof homogenization to be used.
Installation of pressure cells
As part of the backfill of the homogenizedsubsoil, pressure cells (Geokon, Model 4800 and4810) were continuously fitted, successively in two
depth horizons (-0.800m, -0.350m) below the future
foundation base and in the foundation base (0.000m).
344

GEOMATE- Bangkok, Nov. 14-16, 2016

The cells were fitted at three locations in each of thehorizons: the centre, the centre of the edge andcorner of the experimental foundation slab. The cells
in the centre of the edge and corner of thefoundation slab were placed 5cm from the edge ofthe foundation slab, i.e. 0.835m from the centre of
the slab. A total of 9 pressure cells in three horizons
and three vertical axes were fitted (fig. 4). The
placement of the pressure cells in various depth
horizons was derived by comparing the nominal
bearing capacity of the pressure cells used
(2x170kPa, 6x350kPa, 1x700kPa) and the expecteddevelopments of stresses in the subsoil beneath theloaded experimental foundation according to thetheory of elastic half-space defined by Boussinesq.

parameters compared with the original soil. This
could negatively affect the registered stress in thepressure cell. The foundation base was fitted withcells (fat-back pressure cell, type 4810) intended forcontact with the soil and the slab. The side of thecell touching the slab is reinforced with a steel plate
due to specific developments in the stress on theinterface between the soil and the slab. The methodof mounting pressure cells was adapted to the layout
conditions under the foundation of the present
experiment, while respecting the recommendations
made by the pressure cell manufacturer.
Pressure cells intended for the monitoring of thestress in the soil environment record stress, which is
calculated as the sum of total stress (σ) andatmospheric pressure (P). Effective stress (σ') can becalculated using the following Terzaghi’s Eq. (1),provided that the place where the pressure cell islocated is fitted with a piezometer for monitoringpore pressures (u), and total stress (σ) is free fromthe influence of atmospheric pressure:σ' =σ −u(1)

The measured values are affected by numerous
factors distorting the results, some of which can becorrected (from a temperature of the atmosphericpressure). One of the factors which, however, cannot
be affected, is the ratio of strain parameters(stiffness) of the pressure cell and the surrounding
soil environment, which should ideally be identical.
The pressure cell with higher stiffness as comparedto the soil environment registers higher values ofstress which do not correspond to the geostatic stressincluding any incremental stress due to the load.
Moreover, the situation is complicated by thedifferent stiffness of the pressure cell itself, which
has a substantially higher circumferential stiffness
than surface stiffness. Generally, there is a direct
correlation between the thickness of the pressure celland the degree to which pressure is affected due tothe different ratios of elasticity moduli of thepressure cell and the environment. The thinner thepressure cell, the less the measurement is affected byerror due to differences in strain parameters of thecell and the environment.
The technology used to fit the pressure cells into
the soil environment also affects the measurement
significantly. The environment around the pressure
cells must be homogeneous and evenly compacted.
Pressure readings on cells may show lower stressvalues than in the case where the soil environment
near the cell is less compacted. The cell does notregister all stresses due to the non-compacted soil,
which becomes consolidated during loading.
Therefore, stress is transferred to the vicinity of thecell and the stress tensor above the pressure celltakes the form of an arch. Conversely, if theFig. 4 Layout of pressure cells
The thickness of backfill layers was modified asnecessary always to have the pressure cells placed in
an additional groove with a depth of about 100mm
in the freshly compacted layer. The bottom of thegroove was levelled off with a steel scraper, fillingup with soil was not accepted due to the potential
creation of a local site with a different degree ofcompaction. Although the risk of a possible
occurrence of stone or boulder was eliminatedduring homogenization, the site where the pressure
cell was placed was subjected to a subjective
penetration check by a thin steel needle to a depth ofabout 5cm at 9 locations. The fitted pressure cell
was gradually covered by homogenized soil andcontinuously compacted up to the level of thesurface of the currently compacted layer. When
compacting the backfill of the cell, the emphasis wasplaced on keeping a consistent degree of compaction
with the surrounding environment. Backfill withsand was excluded due to the different strain
345

GEOMATE- Bangkok, Nov. 14-16, 2016

immediate vicinity around the cell is compacted
more than the surrounding environment, this leads tostress becoming concentrated in this region;consequently, the cell registers a higher stress which
does not correspond to geostatic stress including any
incremental stress due to the load.

constant pressure during one loading step. As aresult of the above, the pressure in the hydraulic loaddecreased depending on the subsoil being strained.
Therefore, the experimental slab was not subjected
to loading by constant force within one loading step.
During the period from subsoil homogenization
until the actual experiment, the test frame wascovered with a tent which prevented potential
saturation of the subsoil as a result of precipitation.
The groundwater level during the experiment wasalso not encountered; it can be thus assumed that theenvironment was not 100% saturated. The load teston the described experiment was performed around
4 months after the subsoil homogenization. During
this period, dissipation of potential local porepressures occurred and humidity anisotropy withdefining vertical axis was created.

DESCRIPTION OF THE EXPERIMENT
The design of experimental slab was developedby the Department of Building Structures and wasbased on the results of existing experiments [9], [10],[11] and the layout constraints of the test frame
structure. Larger sizes of the experimental basis
would lead to the interaction between the foundationand the subsoil being affected by the existing test
frame foundations. A model foundation slab wasmade of plain concrete of 2x2m and a thickness of
0.2m. The class of the concrete according to thecompressive strength was C35/45. During slab
casting, three samples were made for later analysisof strain parameters (table 1).

Table 2 Load steps of experimental slab
Step
Load
[kN]

1

2

3

4

5

6

7

82

166

244

310

385

471

616

Table 1 Foundation slab properties
f ck,cyl[N/mm2]35

f ck,cube[N/mm2]45

DISCUSSION

Young's modulus [GPa]
29

30,5

The course of the interaction between thefoundation slab and the subsoil is primarily affectedby the dimensions of the foundation and the ratio ofstrain parameters of the foundation and the subsoil.
The experimental foundation slab which, at thisstage of research, was not reinforced had thecharacter of a pliable foundation, where unevensettlement can be expected. As a consequence of low
stiffness of the foundation slab, loading was
accompanied by uneven settlement and uplift of thecorners above the ground, and a gradual reduction in
the contact area between the foundation and thesubsoil. Due to the low tensile strength of concrete(approximately 10x less than the compressive
strength), very soon (in the third step) the firsttensile cracks emerged, developing from thefoundation base (at a load of approximately 250kN).
During the subsequent loading step, the cracks
spread over the entire height of the foundation slab,
resulting in a significant reduction of the contact
area between the foundation and the subsoil. The
load of the slab redistributed to a smaller area thenresulted in a significant increase in stress in thisreduced contact area between the foundation and thesubsoil.
The following figures show the results ofmonitoring total vertical pressure on cells located in
the subsoil under the experimental foundation slab
according to figure 4. The pressure distribution
diagrams are already free from the influence of
atmospheric pressure, temperature and stress due tothe weight of the overlying strata, i.e. they are total

28

The load was applied to the foundation slab bymeans of a hydraulic press leaning against the test
frame structure. Loading took place in several steps,
each taken every 0.5 hours. Each loading step was
about 80kN, with a total of 7 steps taken, i.e. the slab
was subjected to a theoretical maximum load of
about 600kNm. These load values could not be
achieved with total precision due to gross dosing of
the press used. The real values of the load inindividual steps and are summarized in Table 2.
Each loading step involved the following steps:
the increase in load ΔF=80kN (the loadincrement was achieved in about 1 minute),
stress readings on pressure cells madeimmediately after loading (about 3 minutes),stress readings on pressure cells madebefore terminating the loading step (about 3
minutes),the increase in load in the next step.
The experimental foundation was loaded by the
proposed scheme regardless of any defects occurring
during loading. Continuously with measuring,temperatures were recorded on individual pressure
cells, including atmospheric pressure, which waslater used to calibrate the measured data. Also,
settlement of the foundation slab was recorded using
electronic linear path sensors. Pore pressures were
not recorded.
Hydraulic equipment used by the implementationteam during the experiment did not allow keeping
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GEOMATE- Bangkok, Nov. 14-16, 2016

incremental stresses developed in the subsoil only
due to load. The results are systematically divided
into diagrams according to the depth location of thepressure cells and their horizontal position about thefoundation slab.
Figure 5 (z=0.00m) summarizes the progress instress acting on the three cells located in thefoundation base. The solid line characterizes stressbelow the middle, where a significant increase instress during the third loading step is evident (1.5hours). The cell located under the corner
simultaneously stopped recording stress due to the
uplift of the corners of the experimental slab
described above.

Figure 6 and 7 presents the results of the trio
cells located at a depth of 0.35m and 0.80m,
respectively. Here, the qualitative development ofstress is similar, differing only in absolute pressure
values. Even at these depths, a sharp rise in stress isevident, due to the reduction of the foundation base.
The red dot in diagrams presents the stress forloading step before the rupture of the foundationslab, determined based on the theory of elastic halfspace (Boussinesq).
CONCLUSION
The paper aimed to present the results of aphysical model, including preparatory work
involvingthepreparationofsubsoil(homogenization) and mounting pressure cells tomonitor total stress. In particular, stressredistribution in the subsoil below the loaded
experimental foundation slab was monitored. During
these activities, several facts have been confirmed,
which should be summarized in the conclusion.
The importance of the correct positioning of thepressure cell in the soil environment proved to be
crucial. It is essential that material and statehomogeneity of the immediate environment aroundthe cell is observed when the pressure cells are being
fitted. Failure to observe this will significantly affectthe monitoring of total stress in the subsoil. The ratioof strain parameters of pressure cells and subsoil
also has a significant influence. Generally, the stiffer
the cell compared to the soil environment, the higher
total stress it reads. The above information must be
observed when fitting pressure cells as well as wheninterpreting the readings.
The readings of incremental pressures in theselected loading step were evaluated and comparedwith the readings based on the elastic half-space
theory (Boussinesq). The above comparison was
performed using the loading step before thedisruption of the experimental foundation slab. Afterthe disruption of the experimental basis, the contact
area at the foundation base was reduced, which
greatly affected the interaction between thefoundation slab and the subsoil. It should be borne inmind that the theory does not allow taking accountof the stiffness of the foundation slab, whichsignificantly affects the settlement and theredistribution of stress in the subsoil. The
comparison must, therefore, be considered asindicative.
In the course of the experiment, there were a few
questions that will need to be answered during
further research in this area. The physical model will
be extended to include the possible influence ofgroundwater level with a possible modification of its
depth. For these purposes, it will be necessary toincrease the permeability of the subsoil. To increase
the permeability by e.g. mixing the original soil with

Fig. 5 Total pressures in foundation base duringloading

Fig. 6 Total pressures in horizon -0.35 duringloading

Fig. 7 Total pressures in horizon -0.80 duringloading347

GEOMATE- Bangkok, Nov. 14-16, 2016

sand, it would be necessary to add an amount ofsand larger than the volume of the original soil,
which would inter alia, increase the strain
parameters of the subsoil. This is not desirable due
to changes in relative stiffness in the foundation–
subsoil system. The effort to increase the settlementof the experimental foundation as a result of a load
applied to it could, therefore, lead to the bearing
capacity of the test frame being exceeded. Anotherproblem involves the fact that controlled saturationof the subsoil is time-consuming and that it isaffected by capillary elevation, which is problematic.
One possible solution to that situation is to create asandwich of layers of the original soil (approx.
0.20m thick) and a drainage layer (approx. 0.05m
thick). Any internal erosion of the original soil isprevented by inserting a non-woven geotextile
beneath the drainage layer. Such a layer structure
would allow changing the groundwater level in
0.25m increments in a relatively short period thanks
to a significantly shortened drainage path, while
preserving the strain parameters of the original
subsoil.

Politecnica delle Marche, Associazione
Geotecnica Italiana, ISSMGE, 2007.
[4] La Gatta D. P., Keller T. O. “Vertical Load
Tests of Footings on Silt“, First International
Conference on Case Histories in GeotechnicalEngineering, Missouri University of Scienceand Technology.
[5] Anderson J. B., Townsend F. C., Rahelison L.
“Load Testing and Settlement Prediction of
Shallow Foundation“, Journal of Geotechnicaland Geoenvironmental Engineering, American
Society of Civil Engineers, 2007, Vol. 133,
Issue. 12, pp. 1494-1502. ISSN: 1943-5606.
[6] Cajka R., Buchta V., Burkovic K., Fojtik R.
“Experimental Measurement of Ground Base
Plate“ Transactions of the VŠB - Technical
University of Ostrava, Civil Engineering Series,
2012, Vol. 12, Issue 2, pp. 1-6, ISSN: 18044824 (Online), DOI: 10.2478/v10160-0120014-6.
[7] Stolarik, M., Hrubesova, E., Pinka, "M.
Dynamic Response of the Experimental
Foundation Slab - Seismic Measurement andAnalysis". 15 th GeoConference on Scienceand Technologies in Geology, Exploration and
Mining, SGEM International Multidisciplinary
Scientific GeoConferences, Vol. 2, no. 1, p.
261-268. Albena, 2015.
[8] Cajka R., Krivy V., Sekanina D. “Design andDevelopment of a Testing Device forExperimental Measurements of Foundation
Slabs on the Subsoil“, Transactions of the VŠB- Technical University of Ostrava, Civil
Engineering Series, 2011, Vol. 11, Issue 1, pp.
1-5, ISSN: 1804-4824 (Online), DOI:
10.2478/v10160-011-0002-2.
[9] Cajka, R., Labudkova, J., Mynarcik, P.
“Numerical solution of soil - foundationinteraction and comparison of results withexperimetal measurements“, International
Journal of GEOMATE, vol. 11, no. 1, pp.
2116-2122.
[10] Lahuta, H., Hrubesova, E., Duris, L., Petrasova,
T. “Behaviour Subsoil of Slab Foundation
under Loading“ 15th Geoconference on
Science and Technologies in Geology,
Exploration and Mining, SGEM International
Multidisciplinary Scientific GeoConferences,
Albena, 2015. Vol. 2, no. 1, pp. 119-126.
ISSN: 1314-2704.
[11] Cajka, R., Mynarcik, P., Labudkova, J.
“Experimetal measurement of soil-prestressedfoundation interaction“, International Journalof GEOMATE, vol. 10, no. 4, pp. 2101-2108.

ACKNOWLEDGMENTS
This outcome has been achieved with the financial
support of the project GACR No. 16-08937S „Stateof stress and strain of fiber reinforced composites ininteraction with the soil environment“. In thisundertaking, theoretical results gained by theconceptual development of research, development,
and innovations for 2015 at the VŠB-Technical
University of Ostrava (granted by the Ministry of
Education, Youths and Sports of the Czech
Republic) were partially exploited.
REFERENCES[1] Milev N., Kostov V. “Monitoring of thefoundation of an existing RC building andsomemodelling
approaches“,
NZSEE
Conference 2015, Wellington, New Zealand,
pp. 199-206.
[2] Mendoza M. J., Rufiar M., Ibarra E., Mendoza
S.A. “Performance of a pioneer foundation of
the skirt type for the Metro-Line 12 overpass“,Proceedings of the 18th International
Conferenceon
Soil
Mechanics
andGeotechnical Engineering, Paris 2013.
[3] Sokolic I., Ivsic T. “Analysis andmeasurements of settlement for heavy loaded
rigid footing“, 8th European Young
GeotechnicalEngineers'
Conference/
Fratalocchi, Evelina (ur.). - Ancona: Universita

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

RELATING TRADITIONAL DYNAMIC SHEAR MODULUS TONANOINDENTATION MODULUS OF ASPHALT BINDERS
Rafiqul A. Tarefder1, Hasan M. Faisal2, Umme A. Mannan2
Professor of Civil Engineering, University of New Mexico, USA
2
PhD Candidate and Graduate Research Assistant, University of New Mexico, USA
1

ABSTRACT
Traditionally, asphalt binder is characterized by a dynamic shear rheometer, which applies a shear load on a bulk
volume of liquid asphalt binder to determine shear modulus (G*). Recently, a nanoindentation test cancharacterize an asphalt binder film in the form of a coating around roadway aggregates, which is more practical.
In a nanoindentation test, a sharp tip is used to indent an asphalt film while residing on an aggregate surface todetermine nanoindentation modulus (E). This study evaluates whether there is a relation between G* and E. For
both tests, replicate samples were conditioned in three ways: unaged, rolling thin film oven aged, and pressure
vessel aging. Results show that the E-value is approximately 2 to 6 times larger than the G*-value based on all
samples/conditioning.
Keywords: Nanoindentation, Asphalt, Aging, Modulus, Binder, Dynamic, Shear.
INTRODUCTION

THEORY OF NANOINDENTATION

Asphalt concrete’s dynamic modulus (E*) is
one of the key input the parameters for the structuraldesign of flexible pavement according to the
Mechanistic-Empirical Pavement Design Guide(MEPDG, currently packaged as DARWin-ME
software). Currently, asphalt binder shear modulus
G*-based Witczak equation is used in MEPDG
Level 2 analysis to estimate E* of asphalt concrete
from Hot Mix Asphalt (HMA) mix volumetrics andbinder information. Specifically, asphalt binder’sdynamic shear modulus (G*) and phase angle (δ) areused in G*-based Witczak equation to predict E*
mastercurve at Level 2 MEPDG analysis.

A typical nanoindentation load displacement
curve is shown in Fig. 1(a). A sitting load of 0.005
mN is typically applied initially to facilitate acontact between the tip and sample surface. Next,
the load is increased gradually from point a to b. The
tip is unloaded at the maximum load point b. The
unloading path is assumed to be elastic for most of
the elastoplastic material. The unloading curve does
not come back to point a due to plastic deformationin elastoplatic materials. The slope of the unloading
curve at point b is usually equal to the slope ofloading curve at point a.
Fig. 1(b) shows the surface profile as a functionof penetration depth during loading and unloading.
Here, h max is the total depth of indentation at a
maximum load, h p is the total depth of indentationthat is unrecovered, h s is the depth of the surface atthe perimeter of the indenter contact and h c is thevertical depth along which contact is made between
the indenter and the sample. Therefore,

While Asphalt Concrete’s (AC) E* testing is
conducted using an axial compression (or tension),asphalt binder’s G* is conducted in shear mode.
Because, over the years, test methods developed andperformed to characterize asphalt binders have been
limited to a few rheological tests. Moreover, theexisting tests cannot be performed on asphalt binder
film while they are integral parts of AC. Rather
those tests are performed on bulk liquid asphalt
separately. Nanoindentation has created an
opportunity to determine the stiffness (E) of asphalt
binder while they are parts of AC and undercompression loading [].

(1)
hc = hmax − hs
The depth of impression that is recovered is,
he = hmax − h p
Hertz found the contact radius a is related to theindenter radius R, applied load P and the reduced
elastic modulus E* of a sample by (Fig. 2):
3 PR
a3 =
4 E*

To this end, this study conducts dynamic shear
modulus (G*) and nanoindentation stiffness (E)testing of selected Performance Grade (PG) asphalt
binders of New Mexico sources. In addition, thestudy examines possible correlation between G* and
E. It is expected that asphalt binder’s E will a better
estimator of asphalt concrete’s E* for Level 2
MEPDG analysis.

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(2)

(3)

GEOMATE- Bangkok, Nov. 14-16, 2016

Slope

Load

hc = vertical depth alongb which contact is made;he = elastic depth
recovery during
unloading;
hp = Final depth after
unloading;hmax = Depth at maximum
load
Slope hs = displacement of thesurface at the perimeterof the contact.

a

hs
hp
he
hc
hmax

E* =

InitialSurfaceP

Final Surface

(6)

(7)

(8)

(9)

where h is the depth of penetration,𝐡𝐡𝐟𝐟 is plastic depth,
𝛂𝛂 and 𝐦𝐦 are curve fitting parameters related to tip
geometry.
m = 1 for flat ended cylindrical tip, m = 1.5 forspherical tip, and m = 2 for conical tip (Berkovichtip). The slope is measured by differentiation the inabove Eq. (9) at the onset of unloading.

hp

Total
depth,
hmax

π

(S )
2 A
Oliver and Pharr (1992) used a power law function
to fit the unloading path of the load-displacement
curve [5]. The power law function used by OliverPharr is shown in Eq. (9):P = α (h − h f ) m(a) Load –displacement curve

Load

Determination of E*
By differentiating Eq. (5) with respect to thepenetration depth h. Using the relation in Eq. (4):dP
= 2 E * Rh = 2 E * adh
The projected area at the maximum load can be
defined as: A =πa2
Therefore,
dP
2
S=
=
E* A
dh
π
where S is the unloading stiffness or slope of theunloading curve; Therefore,

DisplacementcFig. 2. Indenter Geometry.

hs

Determination of A
hc = hmax - hs

Surface profile
under maximum
load

Oliver and Pharr (1992) defined the projectedarea A as a function of h c defined in Eq. (1) [5].
Oliver and Pharr (1992) extrapolated the tangent line
to the unloading curve at the maximum loading
point down to zero load. This yields an intercept
value for depth which estimates the h s by:P
hs = ε max
S
Therefore,P(4)
hc = hmax − ε max
S
where 𝛆𝛆 is a geometric constant.
ε = 0.72 for conical tip, ε = 0.75 for Berkovich tip,(5)and ε = 0.72 for spherical tip.

(b) Indentation Depth
Fig. 1. Schematic of Indentation Test.
Contact radius a is also related to the indenterradius R and penetration depth by:

a = RhFrom Eq. (2) and (3) the applied load can be writtenas:4
P = E * R1 / 2 h 3 / 2
3

350

(10)

(11)

GEOMATE- Bangkok, Nov. 14-16, 20162.10 MPa for 20 hours according to the AASHTO R
28-12. Fig. 3 shows the PAV apparatus.

Determination of E
Timoshenko and Goodier found the reduced
elastic modulus, E* is related to the modulus of theindenter and the specimen and given by:1
1 −ν 1 −ν i
=
+
E*
E
Ei
where E is Young’s modulus of the material,𝝑𝝑 is Poisson’s ratio of the material,
𝑬𝑬𝒊𝒊 is Young’s modulus of the indenter and𝝑𝝑𝒊𝒊 is Poisson’s ratio of the indenter,
𝑬𝑬∗ is the reduced modulus. One can find the elastic
modulus of the sample, E using Eq. (12).

(12)

Determination of H
Hardness, H, is defined by the maximum load
divided by the projected area (3):P
H = maxA
where P max = peak load and A = projected area ofcontact at the peak load. The unit of hardness isgiven in N/m2=Pa.

Fig. 3. Pressure Aging Vessel (PAV) for long termaging simulation on asphalt binder.
(13)
Nanoindentation Sample Preparation
Fig. 4 shows a laboratory prepared asphalt
binder film on glass substrate. As the first step, a
glass slide surface 0.5 in × 0.5 in was selected andweighed in scale up to 4 significant decimal digits ofgrams. Next, the glass slide was wrapped with hightemperature resistant tape. The tape was placed sothat it formed the 0.25 inch square gap area
previously outlined for the binder. Then, hot
polymer modified liquid asphalt binder was poured
into the gap between the tape strips. The polymermodified binders were melted by heating them to
163oC for an hour. The asphalt coated surface wasplaced in the oven at 163oC for 10 min in order tohave a smooth surface, cooled at room temperatureand the tapes were removed. Finally, the glass slidewith the asphalt coating was weighed again tomeasure the amount of asphalt binder. From the
known area, the density of the asphalt binder andmass the thickness of the binder film was measured.
The film thickness varied within a range of 40 µm to80 µm to avoid the substrate effect on test results.

MATERIALS AND SAMPLE PREPARATION
Collection
A total of four performance grade (PG) binders
available in New Mexico were collected in five onegallon buckets per PG binder.
Table 1: Asphalt binders collected so far
PG 58-22

PG 64-22

PG 70-22

PG 76-22

Some performance graded asphalt binder such as PG
58-34, PG 64-34, PG 70-34 are not used in New
Mexico. These binders could not be collected.
Conditioning
One PG binders were subjected to three
conditioning – (1) unaged/ original binder and (2)RTFO (Rolling Thin Film Oven) and (3) PAV(Pressure Aging Vessel) aged before DSR testing.
The conditioned binder is PG 76-22. The RollingThin Film Oven (RTFO) Test simulates the shortterm aging of asphalt binders that occurs during the
hot-mixing process. Asphalt binder was exposed toan elevated temperature to simulate manufacturing
and placement aging according to AASHTO T 24009 (2009) test procedure. Unaged asphalt binder was
taken in cylindrical glass bottles and placed in arotating carriage within an oven which rotates within
the oven at 325°F (163°C) temperature for 85
minutes. After the RTFO aging, 50gm sample wasplaced in the PAV plate and then aged at 100°C and

Fig. 4. Asphalt Binder Sample for Nanoindentation.

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GEOMATE- Bangkok, Nov. 14-16, 2016

Tarefder and Faisal [1]. The indentation depth
remained small compared to the total material
thickness so that the substrate effect on determining
the mechanical property of the material could be
avoided. The indenter moved at a rate of 1µm/sec tomake the initial contact. In all the tests, the camber
test temperature was kept at 26˚C, within afluctuation of ±0.2˚C. After the test, the temperaturecorrections were also provided to the analysis. In thetest, for each sample a set of 5 indentations (see Fig.
6) were made on the sample with a distance of
300µm. The distance was selected to avoid the pileup and sink in effect for successive indentations.
However, according to ASTM guidelines, therequired distance needed to be at least six indent
radii away from the previous indentation point [6].
Because of the soft bulk of the asphalt binder, thepile up effect could be more. Furthermore, there wasno limitation of space in the sample in nanoscale.
For these reasons, a substantial distance was chosen
for testing the material. Three different aging
condition of the samples of the asphalt binder wereintroduced 20 indentations were conducted for each.

TESTING
Dynamic Shear Rheometer (DSR) Testing
Dynamic Shear Rheometer (DSR) was conducted on
the conditioned binders. The test measures the
dynamic shear modulus or complex modulus, G*
and phase angle (δ) of asphalt binders under acontinuous sinusoidal loading using a dynamic shear
rheometer and parallel plate test geometry. Fig. 5
shows the DSR test device. The G* and δ were
measured at room temperatures and a loading
frequency of 10 rad/sec. A thicker sample (2 mm)with a smaller diameter plate (8 mm) was used sothat the phase angle (δ) becomes measurable. All the
tests were conducted following the AASHT0 T 31509 (2009) test protocol within the linear viscoelastic
range under strain-controlled mode. Three replicate
samples were tested and the average of the threetests was used to develop the master curve.

Fig. 6. Nanoindentation on Unaged Sample.
Fig. 5. Dynamic Shear Rheometer.

RESULTS AND DISCUSSIONS

Nanoindentation Testing

In the current study a comparative analysis has been
conducted between complex modulus of the binder
G* and nanoindentation modulus E of the material.

Indentation experiments were performed by using ananoindenter manufactured by MicroMaterials Ltd.
Wrexham, UK. In all the tests the nanoindenter was
equipped with the pyramidal Berkovich tip. The
indentation tests were conducted in load control
mode. In load control mode, the indentation includes
a constant loading, unloading rate and a holding
segment at maximum load. A maximum load of
0.055 mN was applied with an initial load of 0.005
mN. Thirty indentations for loading rate 0.002
mN/sec, unloading rate 0.002 mN/sec and dwell
time of 200 sec is selected. The loading, unloading
rate and dwell times were selected so the tests could
be performed in the thin film without hitting theglass substrate and also according to the study of

Figure 7 shows the Complex modulus G* offour different binders and three different aging
conditions. The figure shows with an increase inaging the complex modulus of the binder increases,
however the increase is comparatively higher for
PAV aging condition. Since PAV aging condition
simulates long term aging in asphalt binder, thecomplex modulus is almost 3 to 5 times higher for
PAV aging condition, however that is for RAFO
aged condition is only 1.1 to 1.9 times higher.

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GEOMATE- Bangkok, Nov. 14-16, 2016

However the increase in the factor n (=E/G*) shows
decreasing trend for PAV aging condition of PG 5822 and PG 70-28. E might have been showed higher
values compared to the G* of the material.

8.00
UnagedRTFO

PAV

G* (MPa)

6.00
4.00

30.00

2.00

25.00

Unaged

RTFO

PAV

E (MPa)

20.00

0.00

15.00

PG 58-22 PG 64-22 PG 70-22 PG 70-28
PG Binder GradeFig. 7. G* of different PG grades and two different
aging conditions

10.00
5.00
0.00

Figure 8 shows the load displacement curves
obtained from unaged PG 70-22 asphalt binder.
Twenty indentations are conducted on each binder
sample. The load displacement curve shows
maximum displacement is restricted to 7200 nm.

PG 58-22 PG 64-22 PG 70-22 PG 70-28

PG Binder GradeFig. 9. E from nanoindentation test for different PG
binder grade and aging condition
Since both of these binders are manufacture withpolymer modification in the binder industry, thepolymer structure may reduce the effect of aging on
nano scale structure of the binder material.
10
UnagedRTFO

PAVn = E/G*

8
6
4
2

Fig. 8. Load displacement behavior of PG 70-22
unaged asphalt binder.

0
PG 58-22

The selection high unloading rate and extended
dwell time made the unloading portion of the load
displacement curve positively sloped. Therefore, it
was possible to analyze the unloading data in OliverPharr framework. Figure 9 shows the modulus
determined from nanoindentation test. Results show
again with an increase in aging condition themodulus of the material increases. The increase inmodulus is 4 to 8 times higher in the case of RTFO
aged condition, however it is for PAV condition is
10 to 23 times. Therefore, for all cases of thenanoindentation modulus found higher thancompared to complex modulus G* of the material.
Because of the scale different also the surface
energy the nanoindentation. Comparison between E
and G* shows nanoindentation E of the material is
1.5 to 6 times higher than the G* of the material.

PG 64-22

PG 70-22

PG 70-28

PG Binder GradeFig. 10. n-value comparison for different PG binder
grade and aging condition
CONCLUSION
Following conclusions can be drawn from thecurrent study:i) Both Complex modulus and nanoindentation
modulus increases with increase in aging conditionof the material.
ii) Nano scale modulus of the binder is 2 to 6
times higher than micro scale modulus or complex
modulus of the material irrespective of aging
condition.

353

GEOMATE- Bangkok, Nov. 14-16, 2016[3] Tarefder, RA and H. M. Faisal, Modeling
Nanoindentation Creep Behavior of Asphalt
Binder. Advances in Civil Engineering
Materials, ASTM, Vol. 2, No.1, 2013, pp. 418-iii) Nano scale modulus increases 4 to 8 times
for RTFO aging condition where micro scale
modulus shows an increase of 10 to 23 times for
PAV aging condition with respect on unaged asphalt
binder.

440.
[4] Faisal HM, R. A. Tarefder, and M. I. Hossain.

ACKNOWLEDGEMENTS

A Nanomechanical Characterization of Vapor
Condition
andUnconditioned
Asphalt.
Transportation Research Record: Journal of theTransportation Research Board, No. 2506,
Transportation Research Board, Washington,
D.C., 2015, pp. 126–136, DOI: 10.3141/250614.

The authors would like to express their sincere
gratitude and appreciation to the Safety andOperations of Large-Area Rural/Urban Intermodal
Systems (SOLARIS), a tier 1 university
transportation center led by the University of
Nevada, Reno, for providing the funding of thisresearch project.

[5] Oliver, WC, and GM Pharr, An improved
REFERENCES
technique for determining hardness and elastic
modulus using load and displacement sensing
indentation experiments, J. Mater. Res., 1992,
Vol. 7, No.6, pp.1564–1583.

[1] Tarefder RA and H. Faisal, Effects of Dwell
Time and Loading Rate on the Nanoindentation
Behavior of Asphaltic Materials. Journal of
Nanomechanics and Micromechanics, ASCE,
Vol. 3, No. 2, 2013, pp. 17–23.
[2] Tarefder, RA and H. M. Faisal, Nanoindentation
Characterization of Asphalt Concrete Aging.
Journal of Nanomechanics and Micromechanics,
ASCE, 2013, Vol.4, No.1, 2013 A4013003.

[6] ASTM Standard E2546 – 07, Standard Practicefor Instrumented Indentation Testing, ASTM
International, West Conshohocken, PA, DOI:
10.1520/E2546-07, www.astm.org.

354

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

SLOPE FAILURE MODEL TESTS CONTROLLED AGAINSTTHE FAILURES BY USING A FILTERING MATERIAL AND
THEIR SIMULATIONS
Anusron CHUEASAMAT1, Toshikazu HORI2, Hirotaka SAITO1 and Yuji KOHGO1
United Graduate School of Agricultural Science, Tokyo University of Agriculture and Technology, Japan
2
National Institute for Rural Engineering, Japan

1

ABSTRACTThe slope failures caused by heavy rainfall have frequently occurred in many countries. In order toinvestigate the mechanism of the failures, we conducted a series of slope failure model tests. It was found fromthe model tests that the slope failures initiated at the toes of the slopes where the rainwaters infiltrated wereaccumulated and the pore water pressures changed to be positive or be larger than the air entry value of thesurface soil. In order to control the slope failures, drainage of the accumulated rainwaters is important. In thispaper, a filtering material filled within gabions was used to drain the rainwaters. Four model test cases with thedifferent sizes of gabions were carried out. It was found from the results that the slope failures highly depend on
the ratios of the heights between gabions and the slopes. The higher ratio, it took the more time to failure. The
failures always only occurred above the gabions. The simulations of the tests were conducted by using thesaturated and unsaturated consolidation method with an elastoplastic model considering two suction effects. The
simulation results could well express the experimental ones.
Keywords: Slope Failure, Pore Water Pressure, Filtering Material, Saturated and Unsaturated Consolidation
Method, Elastoplastic Model with Two Suction Effects

parts: an artificial rainfall instrument and an iron soiltank shown in Fig. 1. The artificial rainfall
instrument was mounted about 4 meters above thesoil tank to supply rainwater as uniformly as
possible. Square spray nozzles (IKEUCHI Co., Ltd)
were used to supply rainwater. These nozzles could
supply uniform rainfalls to the slopes and the size ofeach rain drop was so small that it had little
influence to slope failures. Some calibration rainfall
tests were conducted to adjust the nozzle layouts. As
the result, we could obtain 49.95 mm/h for the target
50 mm/h rainfall intensity.

INTRODUCTION
The slope failures caused by heavy rainfall havefrequently occurred in many countries (e.g. Tohari etal., 2007). Most of the failure slopes had following
common features: the depths of the failures wereless than 2m, the inclines of the slopes were 30–50
degrees and the slopes consisted of permeable
residual layers on relatively firm rock foundations(Rahardjo et al., 2005; Orense et al., 2006). It was
found from some model failure tests, which
simulated typical natural slopes, that the slope
failures initiated at the toes of the slopes where therainwaters infiltrated were accumulated and the porewater pressures changed to be positive (Orense etal., 2004; Tokoro et al., 2012). In order to control theslope failures, drainage of the accumulated
rainwaters at the toes is important.
In this paper, a filter material filled withingabions was used to drain the accumulated
rainwaters at the toe. Gabions are usually used tostabilize shorelines, stream banks or slopes
against erosion. In this paper, in order to investigate
the drainage performance of the gabions, four
different sizes of filter gabions were used.
EXPERIMENTAL METHODS

Fig. 1 Slope model test framework

The slope model test framework consists of two

355

GEOMATE- Bangkok, Nov. 14-16, 2016

(42,120)

Kasumigaura sandP2
(42,105)

P5
P6
P7(67,95)(67,87)(67,80)P3

P11(92,70)(42,70)P12

P8(67,60)(92,62)

P13(92,54) P16

DL clayP9

P14(67,45)(112,50)(92,45) P17

P18
P4

P15

P10
(42,30)(67,30)Gabion(112,42)P19(112,34) (127,34)(92,30)P20(127,25)yx

Unit: cm

Fig. 4 Pore water pressure transducer positions

Table 1 Initial condition of slope models

Fig. 2 Slope model dimensions and shapes
The slope models used in this series of tests areshown in Fig. 2. The models consist of threematerials: Kasumigaura sand for the surface layers,
DL clay for the firm foundations of the slopes and afilter material within gabion parts. Each slope model
had a shape where a shallow surface sand layer lay
on a relatively hard foundation like typical naturalslopes. The grading curves of three materials areshown in Fig. 3 and all grading curves are almost
parallel. Each slope model was constructed in a steelsoil tank with a transparent glass surface in the frontof the tank. The dimensions of the model slopes
were 122 cm in height (including the 100 cm height
DL clay and the 22 cm height Kasumigaura sand),42 cm in crest width, 70 cm in depth, and 184 cm inbase length. The gradients of slopes were 45 degrees.
Four different sizes of gabions were employed. The
difference of the sizes is expressed in the height ratio
h/H, where h is a height of the gabion and H is aheight of the surface slope, respectively.
At first, the base foundation was constructed bycompacting DL clay in the maximum dry density,
namely compaction ratio D = 100 %, and the watercontent w = about 17 %. The thickness of each DL

clay layer after the compaction was 5 cm and thenecessary number of layers until the completion was
20 layers. After the construction of DL clay layer,
the sand layer was laid on the DL clay foundation.
Relative densities D r = 25 % of the sand layers were
employed with the water content w = about 10 %.
The rolled thickness of each sand layer was 4 cm,
and 23 layers were constructed. The filter material
was filled in the gabions in a minimum density:
about 1.60 g/cm3 and each gabion lay on the toe partof the sand layer as shown in Fig. 2. The sizes ofgabions and the initial conditions of models areshown in Table 1.
During the tests, pore water pressures anddisplacements were measured. The pore waterpressures (PWPs) were monitored with 20 porewater pressure transducers (Capacity = 150 kPa).
Each transducer was connected to each ceramic cupwith a thin tube (Outer diameter = about 3 mm)
filled with de-aired water. The dimensions of eachceramic cup were 12.5 mm in diameter and 25 mm
in length and the air entry value was about 200 kPa.
The ceramic cups were saturated with de-aired water
under a vacuum. The 20 ceramic cups were installed
into each soil slope. The installed positions (P1 –P20) are shown in Fig. 4 and the install level of eachceramic cup was the same as that of each associatedtransducer mounted on the back board of the soiltank. To measure the displacements of the slopes,
pictures were taken through the transparent glass ofthe soil tank with a digital camera. The
displacements were calculated by using Particle

Fig. 3 Grading curves of the slope materials
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GEOMATE- Bangkok, Nov. 14-16, 2016

Image Velocimetry (PIV) analysis method (White etal., 2003).
SIMULATION PROCEDURES
We conducted simulations of the model tests justmentioned before to clarify the mechanism of theslope failure. In simulation process, compression
stresses assume to be negative and the following
three assumptions are assumed to be valid: (i) The
soil consists of soil particles, pore water and pore air,(ii) Pore air is always in contact with the atmosphere,(iii) The permeability of pore air is much higher thanpore water. According to these assumptions, thefollowing two field equations in weak forms arevalid:0,∫ (σ ′ + d u + γ F )d u dV =0.
∫ ( q − a + S ε )d u dV =
V

V

ij, j

i,i

ij eq, j

st

*
i

i

r ii

*
w

(1)

considered. In this analysis, an elastoplastic modelwith two suction effects (Kohgo et al., 1993a, 1993b,
2007) was adopted as a stress-strain relationship.
The two suction effects are that (i) an increase insuction increases effective stresses, and (ii) an
increase in suction enhances yield stresses andaffects the resistance to plastic deformations.
The first suction effect can be evaluated bydescribing the relationship between suction andshear strength at the critical state as follows:

σ ij′ = σ ij - δ ijueq ,
ueq = ua - sa=ueq ua -  se + * e s* 
s + ae 
s = ua - uw ,s = s - se .
*(2)

−2.3(1 + e0 )

κ

+ Ki

G=
Gi + γ J J 2 − γ p p ′ ,( s ≤ se ) ,(6)( s > se ) ,

(7)
(8)
(9)

Where σ ij = total stress tensor, u a = pore air pressure,u w = pore water pressure, s = matric suction, s* =effective suction, s e = air entry suction, a e = amaterial parameter, the brackets < > denote a
Heaviside function: the operation < z > = 0 at z < 0
and < z > = z at z ≥ 0.
The second suction effect can be formulated byevaluating the state surface, which expresses
elastoplastic volume changes behavior in
unsaturated soils (Kohgo et al., 1993b). The state
surface can be plotted in the space with the axes:
mean effective stress p’, effective suction s*, andvoid ratio e. The shapes of state surfaces depend on
the types of soils. Here, when s* m = ∞, state surface
is expressed as follows (Kohgo et al., 2007):

Equation (1) is the force equilibrium equationand the Equation (2) is Richard’s mass conservation
equation of pore water. Here σ´ ij = effective stresstensor, d ij = Kronecker’s delta, u eq = equivalent porepressure, γ = unit weight of soil, F i = componentsof the body force vector, q i = components of therelative displacement velocity vector of water withrespect to soil skeleton, a st = change of storage ofwater due to change of degree of saturation (S r ), ε ii
= volumetric strain of soil skeleton, du* i = incrementof the virtual displacement vector, du w * = incrementof virtual pore water pressure. Subscripts after acomma denote spatial differentiation. A superposed
dot denotes differentiation with respect to time.
Unknown variables in these equations are
displacements of soil skeleton and pore waterpressures. The field equations are introduced to thediscrete system using the finite element method andsolution procedure base on the modified NewtonRaphson method is adopted.
In the elastic range, the elastic moduli are
assumed to be functions of p' and the second
deviator stress invariant J 2 as follows:

=
K

(5)

e=
−l *log ( − p ′) + Γ * ,(10)

λ * s**
λλ= + * f1 * ,(11)s + a1

Γ * = e010 +(Γ − e ) λ .
0
01

*

λ

(12)

Where l=l* in saturation; Γ =Γ* in saturation, l* =the slopes of e - log (- p´) curves in the elastoplastic
range, Γ* = void ratios of the e - log (- p') curves inthe elastoplastic range at p´= unit, s* m , l* f1 , a 1 * ande0 01 = material parameters.

(3)
(4)

Finite Element Meshwhere K = the bulk modulus, G = the shear modulus
and K i , G i , γ J and γ p = material parameters.

Figure 5 shows the finite element mesh (FEmesh) and boundary conditions used in thesimulation. Dimensions of the FE mesh are the sameas the experimental slope models. The FE meshconsists of 230 elements and 759 nodal points. The
8-node isotropic elements were used for the
displacement field and super-parametric elements

An Elastoplastic Model for Unsaturated Soil
In conducting consolidation analysis, material
nonlinear properties: stress-strain relationship,
permeability, soil water retention curves have to be357

GEOMATE- Bangkok, Nov. 14-16, 2016

were used for the pore water pressure filed. At first,
built-up analyses were carried out from 1 layer to 14
layers, then, consolidation terms were set and afterthat rainfalls were subjected to the surface of theslope as constant flux values. The value was 50 mm/
h. At the bottom of the model slope, both directionsof displacements were fixed and impervious
condition was set. On the left and right side surfacesof the model slope, only the horizontal direction of
the displacement was fixed. The material parametersof permeability model, SWR model, and otherparameters are shown in Table 2.

positive PWP values appear in the surface layers. In
Case 1, the positive PWP values were seen at only
the toe and the boundary between the sand layer andthe foundation. Meanwhile, in Case 4 the positive
pore water pressure values were monitored in almost
the all areas of the surface layer. Thus the larger sizeof the filter gabions, the more rainwater could be
stored within the slopes. Times t f when the first
failures occurred were 73 min in Case 1, 99 min in
Case 2, 115 min in Case 3 and 148 min in Case 4,
respectively. The times increased with an increase in
the height ratio h/H (see Fig. 7).
Figure 7 shows the relationship between the first
failure occurrence time t f and the height ratio h/H.
The relationship was linear and t f increased with anincrease in h/H. From the figure, the value of t f in
the case without the filter gabion could be read 63
minutes that is almost the same as 55 minutes
obtained from our previous experimental test.
Figure 8 shows displacement vectors during
failures. The large displacements could be only seen
within the surface layer. Especially thedisplacements concentrated on the failure parts thatstarted from the almost the same levels as those of
the tops of filter gabions and extended to the upperparts. The areas where the large displacements weremonitored became smaller in order of the heightratios. In all the cases, the displacements occurredalmost along the slopes.

Fig. 5 Finite element meshes and boundary
conditions used for the model simulation

Table 2 Material parameters of slope model
simulation

Fig. 6 Pore water pressure distributions when thefirst failures occurred
Figure 9 shows pictures at the final situations
taken from diagonally above the slopes. In all thecases, the surface failures appeared. Types of slope
failures depend on the cases. In Case 1, relatively
large and clear failures could be seen. In other cases,
the failure areas became mud and the failures werelike flows.As mentioned before, the failures started
from the almost the same levels as those of the topsof filter gabions and extented to the upper part. The
sizes of failure areas were smaller as the height ratio
increased. Thus, the filter gabions could make not

RESULTS
Experimental Results
The experimental results are shown in Figs. 6 – 9.
Figure 6 shows pore water pressure distributions
when the first failures occurred. In all the cases,
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GEOMATE- Bangkok, Nov. 14-16, 2016

only the failure size smaller but also the duration
time of rainfall until the failure longer. But they did
not express the drainage performance.

Comparing the results at the same elapsed time t =60 minutes, in both positive PWP values appearedalmost behind the gabion. In other parts there were
negative PWP values. In DL clay layer most of areas
were smaller than -10 kPa that was the air entry
value of DL clay. At t = 100 mininutes, there was aclear seepage surface (u w = 0 kPa on this surface)
and the surface appeared at a relatively upper
position of the slope surfaces. The seepage levelcalculated was slightly higher than that measured.
Figure 11 shows a comparison of simulation andexperimental results of displacement vectors in Case
2. It was found that large displacements were
calculated or measured within the surface layer,
especially they were remarkable at the toe of theslope. The both situations were almost similar. At t
= 99 minutes, the first failure occurred at the toe
within the surface layer and the failures graduallyprogressed to the upper parts of the slope. At t = 120
minutes, they advanced to the whole surface slope.
Figure 12 shows the simulation result of shear
strain distributions. At t = 80, 100 minutes, shear
strains were still small. At t = 120 minutes, large
shear strains were calculated within the surface layer
and the area where large shear strains calculated
s
t
a
r
t
e
d

Fig. 7 First failure times versus the height ratios

Fig. 8 Slope displacement vectors during failures

Fig. 10 Pore water pressure distributions in Case 2

Fig. 9 Slope failure patterns at the end of the tests
Simulation Results
A model test: Case 2 was simulated by using asaturated and unsaturated consolidation analysis
method coupled with an elastoplastic model forunsaturated soils (Code name: GEOCUP). The
model had I = 50 mm/h and D r = 25 %. Figure 10
shows the distributions of PWP values in bothexperimental and simulation results for Case 2.

Fig. 11 Slope displacement vectors in both resultscomparison of Case 2359

GEOMATE- Bangkok, Nov. 14-16, 2016

ACKNOWLEDGEMENT
The research herein was supported by the Japan
Society for the Promotion of Science (JSPS) (thegrant No. 23580327). Finally, we would like toexpress gratitude to members of our laboratory andstaffs of National Institute for Rural Engineering,
Tsukuba, Japan for their help and contributions tothis series of experimental model tests.

REFERENCES[1] Kohgo, Y., Nakano, M. and Miyazaki, T.
1993a. Theoretical aspects of constitutive
modeling for unsaturated soils, Soils andFoundations 33(4), 49-63.
[2] Kohgo, Y., Nakano, M. and Miyazaki, T.
1993b. Verification of the generalized
elastoplastic model for unsaturated soils, Soils
and Foundations 33(4), 64-73.
[3] Kohgo, Y., Asano, I. and Hayashida, Y. 2007.
An elastoplastic model for unsaturated rockfills
and its simulations of laboratory tests, Soils andFoundations 47(5), 919-929.
[4] Orense, R. P., Shimoma, S., Maeda, K. andTowhata, I. 2004. Instrumented model slope
failure due to water seepage, J. Natural Disaster
Science 26(1): 15-26.
[5] Orense, R. P., Sapuay, S. E., Billedo, E. B. andMatsuoka, K. 2006. The 2004 sedimentdisasters in Quezon province, Philippines
triggered by heavy rainfall, Soils andFoundations 46(6): 701-711.
[6] Rahardjo, H., Lee, T. T., Leong, E. C., Rezaur
2005. Response of a residual soil slope torainfall, Can. Geotechnical J., 42(2), 340-351.
[7] Tohari, A., Nishigaki, M. and Komatsu, M.
2007. Laboratory rainfall-induced slope failure
with moisture content measurement, J.
Geotechnical and Geoenvironmental Eng.
ASCE 133(5), 575-586.
[8] Tokoro, N., Tanikawa, K., Saito, H., Kohgo, Y.
and Hori, T. 2012. On 1G slope failure model
tests due to rainfalls - Difference of failure
patterns due to difference of densities of a
subsurface sand layer, Proc. 2nd Advances inTransportation Geotechnics: 755-760.
[9] White, D.J., Take, W.A. and Bolton, M.D. 2003.
Soil defor- mation measurement using particle
image velocimetry (PIV) and photogrammetry,
Geotechnique 53(7): 619-631.

Fig. 12 Simulation result: Shear strain distributions
with elapsed time of Case 2
from the same level as that of top of the gabion andextended to the crest of the slope. At t = 140 minutes,
the area where the values of shear strains reachedabout 20 % made a band and the band reached thecrest of the slope. The area where large amounts ofshear strain calculated were almost consistent withthe area where failures were measured from theexperimental test.

CONCLUSION
We conducted a series of slope model tests withdifferent sizes of filter gabions under a heavy rainfall
condition. Each slope model had a shallow and loose
surface sand layer on a relatively hard foundation as
often seen in natural slopes. The filter material filled
in gabions were used to control against slope failures.
A model test was simulated by using a saturated andunsaturated consolidation analysis method coupledwith an elastoplastic model for unsaturated soils(Code name: GEOCUP). We could obtain thefollowing results: (i) Negative pore water pressure
dissipated from the toes of the slopes and thesaturated zones progressed to the upper part of theslopes, (ii) Large displacements were only
accumulated in the surface layer, (iii) The failures
started from the saturated toes at the top of the filter
gabion and the failure zones progressed to upper partof the sand layer, (iv) The larger size of the filter
gabion (high value of h/H ratio) could contain morewater infiltrated inside of the slopes before a failure
occurring, and (v) These situations mentioned above
were consistent with both of the experimental andsimulation results.
The filter gabions could make not only the
failure size smaller but also the duration time ofrainfall until the failure longer. But they did not
express the drainage performance.

360

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

SETTLEMENT PREDICTION OF LARGE-DIAMETER
BORED PILE IN BANGKOK SOILS
Thayanan Boonyarak1, Zaw Zaw Aye1, Nutthapon Thasnanipan1, Sahapap Supawo2, SereyroathChea1 1Seafco Public Company Limited, Thailand; 2Faculty of Engineering, Khon Kaen University,
Thailand

ABSTRACTFor large-diameter pile design, in addition to analysis of allowable bearing capacity, pile settlement
calculation is required for minimizing differential and excessive settlements of a building. The objective of thispaper is to evaluate and compare capability of each approach for pile settlement estimation. Data in this study
was obtained from instrumented static pile load test results on 25 large-diameter bored piles and barrettes
constructed in Bangkok. Thus, pile head settlement, pile base settlement and load distribution in each soil layer
can be investigated. The pile settlement data was back-analyzed using analytical solutions and numericalmodeling. Factors that influenced pile settlement behavior, consisting of soil type at pile base, shape of pile anddrilling slurry, were analyzed. According to measured data, pile constructed using bentonite slurry had larger
settlement than those excavated with polymer slurry. This is because shaft friction reduction in the former waslarger than the latter, resulting in larger load transferred to the pile base and caused higher toe settlement. For
pile settlement more than 2% of pile diameter, axial load transferred to pile toe. In these settlement regions,analytical approaches substantially underestimate the measured results while the numerical modeling stillreasonably captured the measured one. Guidelines for adopting each method for pile settlement calculation are
also proposed.
Keywords: Pile settlement prediction, Load transfer mechanism, Bangkok soils, Load-settlement database,
Numerical modeling

INTRODUCTIONδ =For deep foundation, apart from capacity of pile,accurate settlement calculation is also a key factor tobe considered. To calculate pile settlement,
analytical solution is commonly adopted forsimplicity. Tomlinson (1995) [1] proposed an
equation to calculate pile settlement as shown in Eq.
1. For brevity of the paper, detail of each parameter
is not described. To sum up, the first and second
terms refer to elastic shortening of the pile andsettlement of soil at pile base, respectively.
δ =(𝑊𝑊𝑠𝑠 +2𝑊𝑊𝑏𝑏 )𝐿𝐿
2𝐴𝐴𝑠𝑠 𝐸𝐸𝑝𝑝+

𝜋𝜋𝑊𝑊𝑏𝑏 𝐵𝐵�1−𝜈𝜈2 �𝐼𝐼𝑝𝑝
4𝐴𝐴𝑏𝑏 𝐸𝐸𝑠𝑠

𝑃𝑃

∆𝐿𝐿

𝐴𝐴𝑎𝑎𝑎𝑎 𝐸𝐸𝑝𝑝

(1−𝜈𝜈2 )𝐸𝐸𝑠𝑠

𝑚𝑚𝐼𝐼𝑆𝑆 𝐼𝐼𝐹𝐹 𝐹𝐹1

𝐴𝐴𝑝𝑝 𝐸𝐸𝑝𝑝+

𝑄𝑄𝑤𝑤𝑤𝑤 𝐶𝐶𝑝𝑝
𝐷𝐷𝑞𝑞𝑝𝑝𝑄𝑄+ 𝑤𝑤𝑤𝑤𝐷𝐷(1−𝜈𝜈2 )𝐼𝐼𝑤𝑤𝑤𝑤
𝑝𝑝𝑝𝑝𝐸𝐸𝑠𝑠(3)

Alternatively
fromanalyticalsolutions,numerical analysis can be adopted for computing
pile settlement. Elaziz and Naggar (2014) [4]presented finite element method by back-analyzing
results from hollow steel pile load testing.
Shooshpasha et al. (2013) [5] conducted threedimensional finite element analyses for determining
bearing capacity of a pile and validated withmeasured result.
Construction technique of bored pile also
influences the pile capacity. Thasnanipan et al.
(1998) [6] reported that if the exposure time ofbentonite slurry in each borehole is more than 48
hours without agitating, filter cake thickness is high.
This resulted in substantial reduction in load-bearing
capacity of the pile. To minimize this problem,
Thasnanipan et al. (2003) [7] and Boonyarak et al.
(2015) [8] found that bored pile constructed using
polymer slurry had larger load bearing capacity thanthoseconstructedwithbentoniteslurry.
Submaneewong (1999) [9] studied effect of barretteshape on load bearing capacity and reported
empirical equation to estimate pile behavior in eachtype of soil.

(1)

Bowles (1996) [2] considered pile shortening
according to each layer of soil along pile shaft andcalculate soil settlement at pile toe (refer to Eq. 2).
δ = Σ � 𝑎𝑎𝑎𝑎 � + ∆𝑞𝑞𝑞𝑞

�𝑄𝑄𝑤𝑤𝑤𝑤 + 𝜉𝜉𝑄𝑄𝑤𝑤𝑤𝑤 �𝐿𝐿(2)

Vesic (1977) [3] adopted three components ofpile settlement as stated in Eq. 3. In this equation,
the first, second and third terms are pile settlement
due to elastic shortening, soil compression at pilebase and soil compression along the pile shaft,respectively.
361

GEOMATE- Bangkok, Nov. 14-16, 2016

This paper presents a method for estimating pilesettlement based on load testing database of piles
constructed in Bangkok soils. Factors influenced pilesettlement behavior was analyzed. Limitation andrecommendation of each method is explained anddiscussed.

conventional elastic-perfectly plastic model, thismodel can simulate stiffness dependent on stress,
strain and path. Parameters were adopted andmodified from [14] and [15], who studied andcalibrated the model parameters for Bangkok soils.
All parameters are summarized in Table 1.

BACK-ANALYSIS OF PILE LOAD TESTS

See Fig 1b

P/Ab

80 m

43 m

Data analyzed in this study is obtained from 25
instrumented pile load tests in Bangkok soils. Pile
diameters are ranging from 0.80 m to 1.65 m anddepth of pile tip are varying between 41.0 m and59.3 m. All piles were tested using slow maintained
load test method [10] with maximum test load of 2.5
times the design safe working load.
Analyzed data can be categorized into fourgroups. These groups are circular bored pile withpile tip in sand and clay, barrette with tip embedded
in sand and clay.
For estimation of pile settlement, Young’s
modulus of sand and clay are assumed to be 2000N
and 800C u , respectively (adopted from [11]).
Poisson’s ratio of soil is assumed to be 0.25. For
concrete pile, modulus of elasticity is 27 GPa.
NUMERICAL ANALYSIS OF LOAD TESTING

Pile radius, R = 0.5 m

30 m
(a)

Finite element method using software PLAXIS
2D 2015 was adopted for computing pile settlement.
Fig. 1Finite element mesh and boundary conditions
To model a single pile behavior, axis-symmetry
analysis was carried out using finite element mesh as
shown in Fig. 1a. One project from load testing
database was selected considering the same soil
conditions as load testing. The pile had a diameter of
1.0 m and depth of 43 m.
Boundary conditions at the right side and thebottom of the mesh were roller and hinge,
respectively. The boundary at the left side was theaxis of symmetry. The width of the model is 30 m
and depth is 80 m which is greatly larger than therequired distance of six times the pile diameter at theside and the pile base.
Figure 1b shows details of pile and applied load.
Pile was model using linear elastic material having
Young’s modulus, Poisson’s ratio and unit weight of
27 GPa, 0.15 and 24 kN/m3, respectively. Reduction
factor of 0.75 was applied at the interface between
the pile and the surrounding soil at the side as there
is a slip between the soil and pile. However, nointerface was adopted at the pile base as the soil andpile always attached to each other.

(b)(a) Finite element mesh; (b) Details at pilehead and surrounding soils

Table 1 Parameters for numerical analysis
Para
meter

Medium
stiff
clay

Soft
clay

c'

1

5

Stiff to
very
stiff
clay
11.5

Dense
sand

Hard
clay

0.1

11.5

φ′

27

27

28

33

28

Ψ

0

0

0

5

0

E ref,50
E ref,oe

800

5150

5150

27000

9500

850

6425

6425

27000

12000

E ref,ur

8000

19000

19000

162000

30000

Rf

0.9

0.9

0.9

0.9

0.9

m

1

1

1

0.5

1

K 0, NC

0.74

0.74

0.5

0.5

0.5

ν ur

0.2

0.2

0.2

0.2

0.2

γ 0.7

0.0005

0.001

0.001

0.014

0.0015

G max

5000

16000

48000

200000124000d

Soil model and parameters
Soil was modeled using Hardening soil withsmall strain stiffness [12 & 13]. Unlike other

Modeling procedures
362

GEOMATE- Bangkok, Nov. 14-16, 2016

pile tip in clay was substantially larger than those insand. For example, at P/A s of 60 kN/m2, settlementsof barrette with tip in sand and clay were 0.5%D and
5.5%D, respectively. This observation wasconsistent with those of circular bored pile (refer toFig. 2).
For effects of shape on settlement of barrette, at
δ/D of 2%, P/A s of barrette with tip in sand wasabout 80 kN/m2, which is in the same range of thosefor circular bored pile (constructed using bentonite
slurry) of from 70 to 90 kN/m2. When tip of barrette
was in clay, P/A s was about 50 kN/m2, which issimilar to those of bored pile (50-70 kN/m2). Thisfinding suggests that shape of barrette does not
significantly affect the settlement behavior.

In the first stage, stress in soil is initializedaccording to K0,NC of each soil. After that, bored pileis constructed by changing material properties fromsoil to concrete. It is assumed that there is no soil
displacement during bored pile installation. For pileload testing, line load (P) is applied on cross section
area of pile (Ab). Load is increased by 1% step until
reaching the maximum test load of 100% in each
pile.
INTERPRETATION OF RESULTS
Settlement of circular bored pile
Results of load-settlement relationship of circularbored pile are shown in Fig. 2. For comparison ofdata in each test, applied load is divided by shaft
area (P/A s ) and settlement at pile head is normalized
by its pile diameter (δ/D). General trend of loadsettlement curves can be described by bi-linear
relationship. The first part of linear is where
normalized settlement is within 2%. After that, load
settlement relationship becomes curved or so-called
transition zone. The second part of linear is the zoneof higher settlement rate than the first part of linear.

Normalized pile head settlement (δ/D, %)Applied load / shaft area (P/As, kN/m2)

For bored pile constructed using bentonite slurrywith pile tip in sand, P/A s at normalized settlement
at 2% is between 70 and 90 kN/m2. By comparing
with pile constructed using polymer slurry having
pile tip in sand, the P/A s of this type of pile at thesame δ/D was ranging from 90 to 120 kN/m2. It
suggests that shaft friction of bored pile using
bentonite slurry is smaller than those with polymer
slurry. This is because thicker filter cake in theformer than in the latter [7], resulting in smaller
shaft friction. Note that only one set of data frompile using polymer shows a bi-linear behavior asaxial load did not transfer to the toe of most of pilesof this type.

0

20

40

60

80

100

120

140

160

0
2
4
6
8
10
12

Bentonite, Sand
Bentonite, Clay

Polymer based, Sand
Polymer based, Clay

14

Fig. 2

Load-settlement curves of circular boredpile with different soil types at pile tip anddrilling slurries

Normalized pile head settlement (δ/D, %)Applied load / shaft area (P/As, kN/m2)

The effects of drilling slurry on pile settlement
also influenced the piles with tip in clay. At
settlement of 2%D, P/A s for piles constructed using
bentonite slurry and polymer slurry are 50-70 kN/m2
and 70 kN/m2 (data only from one pile), respectively.
Considering type of soil at pile base, it is foundthat pile with tip in sand had smaller settlement thanthose having tip in clay (given the same type ofdrilling slurry). This is because soil stiffness (e.g.
Young’s modulus) of sand (at depth of more than 40
is substantially larger than clay as stiffnessm)proportionally increases with depth for sand.
Settlement of barrette

0

20

40

60

80

100

120

140

160

0
2
4
6
8
10
12

Measured [Sand]
Calculated [Tomlinson, Sand]
Calculated [Bowles, Sand]
Calculated [Vesic, Sand]

Measured [Clay]
Calculated [Tomlinson, Clay]
Calculated [Bowles, Clay]
Calculated [Vesic, Clay]

14

Fig. 3Comparison of effects of soil type at pile tip
on settlement of barrettes

The three approaches for calculating deep
foundation settlements were compared with themeasured results (refer to Fig. 3). In the region
where pile head settlement was between 1%D and
2%D, the three methods reasonably captured themeasured results as elastic shortening of the pile was
the major factor of pile head settlement. As the load
was increased, settlements calculated by Tomlinson(1995) and Bowles (1996) were significantly
underestimated. The calculated result obtained from

Figure 3 shows load-settlement relationship ofbarrettes with tip in sand and clay. All barrettes wereconstructed using bentonite slurry as polymer slurry
was not applicable to stabilize the trench of barrette.
From the comparison, settlement of barrette with363

GEOMATE- Bangkok, Nov. 14-16, 2016

Vesic (1977) had a larger settlement than the othertwo but still underestimated the measured one. Thisis because shear modulus of soil considered in theanalytical solution was assumed constant, resulting
in underestimation of settlement at higher load. Thiseffect of stiffness on pile settlement is explainedlater in the topic shear modulus degradation
modeling in numerical analysis.

Bowles (1996) and Vesic (1977) were ranging from
1.8%D to 2.0%D. Similarly, pile shortening
obtained from numerical analysis was 1.9%D. It isfound that both approaches provided reasonablevalue of pile shortening compared with the measured
result.
The normalized toe settlement from measured
result was 8.9%D but the calculated results using
analytical solution substantially underestimated themeasured one. However, toe settlement computed
using numerical analysis was closed to the measured
result, resulting in reasonable estimation of pilesettlement calculation (refer to Fig. 4).

Comparison between numerical analysis andanalytical solution

Normalized settlement (δ/D, %)Measured results of pile load test compared withthree analytical solutions and numerical analysis areshown in Fig. 4. This pile had a diameter of 1 m,
depth of 43 m and was constructed by using
bentonite slurry.
From the comparison, analytical solution
acceptably predicted the measured result up to P/A s
of 42 kN/m2 or at 125% of designed safe working
load. When the test load reached its maximum at
P/As of 96 kN/m2, measured settlement was morethan two times larger than those obtained from theanalytical solutions.
Settlement obtained from numerical analysis
slightly overestimated the measured result in thesmall settlement region. At higher testing load, finite
element method result was very similar to themeasured result. Relatively small discrepancy in theregion of P/As from 60 to 80 kN/m2 between
computed and measured one may be due toinconsistent stiffness degradation modeling in the
numerical analysis.
Normalized pile head settlement (δ/D, %)

20

40

60

80

100

120

140

4
2

Proportion of pile settlement

Figure 6 shows normalized shear modulus of soil
(G After /G Before ) at each normalized pile settlement(δ/D). Note that G Before and G After refer to shear
modulus of soil before and during pile load test,
respectively. Shear modulus used in the analyticalsolutions is constant by using an average or lower
bound value. For numerical analysis, the normalized
shear modulus reduced from 1 to 0.34 at pilesettlement of 1.1%D. This settlement is where
friction is almost fully mobilized (between 1-2%D).
As an advanced soil model is adopted in this study,
the effect of soil stiffness degradation can beconsidered.
As shown in Figs. 4 and 5, analytical approaches
substantially underestimate the measured results
while the numerical modeling still reasonably
captured the measured one. This is due to constantsoil stiffness and stiffness degradation adopted forthe former and the latter, respectively.

160

4
6
8
Measured [2540-6-BS]
Calculated [Tomlinson]
Calculated [Bowles]
Calculated [Vesic]
Computed [PLAXIS]

14

Fig. 4

6

Effects of shear modulus degradation of soil onpile settlement2

12

Toe settlement
8

Fig. 50

10Pile shortening

0

Applied load / shaft area (P/As, kN/m2)
010

Measured load settlement curve comparedwith three analytical solutions andnumerical analysis

Proportion of pile settlement
Pile shortening and toe settlement of the pilewhich is discussed in the previous section is shownin Figure 5. Measured result of normalized pile
shortening was 1.4% or equivalent to 0.52
PL p /A b E p ; where P is maximum test load, L p is thepile length, A b is the cross-sectional area of the pileand E p is Young’s modulus of the pile. Pile
shortening calculated from Tomlinson (1995),

ACCURACYOFPILE
SETTLEMENT
ESTIMATION BY ANALYTICAL SOLUTION
Comparisons of maximum measured pile headsettlement and calculated result from the three
analytical solutions are shown in Figure 7.
Figure 7a shows comparison of pile settlement364

GEOMATE- Bangkok, Nov. 14-16, 2016

estimation using Tomlinson (1995)’s approach withmeasured results. At pile head settlement not morethan 2%D, calculated settlement was slightly larger
than the measured one. For this settlement region,
the analytical solution was still reasonably estimatedthe measured settlement with error of from 0.5%D to
1.0%D.
However, for pile with settlement larger than
5%D, the calculated settlement significantly
underestimated the measured result by 3%D to 8%D.
In addition, estimation of pile settlement constructedby using polymer slurry was more accurate thanthose constructed with bentonite slurry and barrette.
Thus, one should be aware when using the analyticalsolution to estimate settlement of pile that has atendency to be larger than 5%.
There are also comparisons between two
analytical solutions of pile head settlement using
Bowles (1996) and Vesic (1977) and the measured
results as shown in Figure 7b and 7c, respectively.
These two analytical solutions provided similar
trend to those obtained from Tomlinson (1995).
Thus, to predict pile settlement, any of the three
analytical solutions can be adopted up to pilesettlement of 5%D.

Calculated normalized pilesettlement, Tomlinson (δ s/D, %)Table 2 summarizes average load per unit area atnormalized settlement of 2% and 5%, which aresettlement at fully mobilized shaft friction andultimatebearing capacity, respectively.
16
1:1

10
8
6
4
2

0

(a)

2

4

6

8

10

12

14

16

Measured normalized pile settlement (δ s/D, %)16
Pile, Bentonite, Sand
Pile, Polymer, Sand
Pile, Bentonite , Clay
Pile ,Polymer,Clay
Barrete,Bentonite,sandBarrete,Bentonite,Clay

Calculated normalized pilesettlement, Bowles (δ s/D, %)

14

0.8

12
10

1:1

8
6
4
2
0

0.6

0

(b)

2

4

6

8

10

12

14

16

Measured normalized pile settlement (δ s/D, %)16

0.4

Pile, Bentonite, Sand
Pile, Polymer, Sand
Pile, Bentonite , Clay
Pile ,Polymer,Clay
Barrete,Bentonite,sandBarrete,Bentonite,Clay

14

Calculated normalized pilesettlement, Vesic (δ s/D, %)Normalized shear modulus(GAfter/GBefore )

12

0

1.0

0.2
Calculated [Tomlinson, Bowles & Vesic]
Computed [PLAXIS]
0.0
0

2

4

6

8

10

12

14

Normalized pile head settlement (δ/D, %)Fig. 6

Pile, Bentonite, Sand
Pile, Polymer, Sand
Pile, Bentonite , Clay
Pile ,Polymer,Clay
Barrete,Bentonite,sandBarrete,Bentonite,Clay

14

Shear modulus degradation and pile headsettlement

OF
LOAD
IMPLEMENTATIONSETTLEMENTDATABASE
FOR
PILE
CAPACITY DETERMINATION

12
10

1:1

8
6
4
2
0
0

According to relationship between applied load
per shaft area (P/A s ) and normalized pile headsettlement (δ/D) as refer to Figures 2 and 3, pile
capacity can be determined. If pile settlement of
5%D is set as an ultimate load bearing capacity
settlement, average P/A s can be obtained for eachtype of pile. For example, bored pile with diameterof 1.5 m, depth of 50 m, embedded in sand andconstructed using polymer slurry is estimated tohave P/A s of 140 kN/m2. Thus, the ultimate load
bearing capacity and a safe working load of this pile
are about 32,000 kN and 12,800 kN, respectively
given that factor safety of 2.5 is adopted.

2

4

6

8

10

12

14

16

(c)

Measured normalized pile settlement (δ s/D, %)Fig. 7Comparison of settlement prediction using
following approach: (a) Tomlinson, 1995;(b) Bowles, 1996; (c) Vesic, 1977

Table 2 Summary of load-settlement relationshipP/A s , kN/m2

365

Pile type

Slurry type

Bored pile, sand

Bentonite

Bored pile, sandPolymer

Bored pile, clay

Bentoniteat δ/D of
2%
70 - 90
90 - 120
50 - 70

at δ/D of
5%
80 - 100
140
60 - 80

GEOMATE- Bangkok, Nov. 14-16, 2016

P/A s , kN/m2
Pile type

Slurry typeat δ/D of2%
Bored pile, clay
Polymer70
Barrette, sand
Bentonite
80
Barrette , clay
Bentonite50
Note: * Pile settlement did not reach 5%D
CONCLUSIONS

at δ/D of
5%
N/A*
90
60

Based on back-analyzed data from 25 numbersof instrumented pile load test, following conclusions
may be drawn:(a) For deep foundation embedded in the same soil
type, settlement of bored pile constructed using
bentonite slurry is substantially larger than those
constructed with polymer slurry. This is becauseof thicker filter cake in former than the latter,causing reduction in shaft friction.
(b) By comparing bored pile constructed using thesame type of slurry, bored piles embedded inclay have larger pile settlement than thoseembedded in sand. This is due to smaller soil
stiffness in the former than the latter.
(c) There is no significant difference in settlement
between barrette and circular bored pile. Thus,
load-settlement relationship of bored pile can be
adopted to estimate the settlement in barrette,given that bentonite slurry is used in bored pileand soil type at pile toe is the same.
(d) Pile settlement estimated using analyticalsolutions proposed by Tomlinson (1995), Bowles(1996) and Vesic (1977) provided reasonable
result in the range where resistance is mainly
provided by shaft friction. This range is withinpile settlement of 2% the pile diameter.
(e) At settlement larger than 5% the pile diameter,
the three analytical solutions substantially
underestimate the measured result. On the other
hand, numerical analysis still reasonably predicts
the measured result. This is because soil stiffness
degradation was modeled in the latter while
stiffness was assumed constant in former.

capacityofhollow-bar“Geotechnicalmicropiles in cohesive soils”, Can. Geotech. J.
51: 1123–1138, 2014, dx.doi.org/10.1139/cgj2013-0408[5] Shooshpasha, I., Hasanzadeh, A. and Taghavi,
A. “Prediction of the axial bearing capacity ofpiles by SPT-based and numerical design
methods”, Int. J. of GEOMATE, June, 2013,
Vol. 4, No. 2 (Sl. No. 8), pp. 560-564[6] Thasnanipan, N., Baskaran, G. & Anwar, M.A.
"Effect of construction time and bentonite
viscosity on shaft capacity of bored piles",Proceeding of the 3rd international geotechnicalseminar on bored and auger piles, Ghent,
Belgium (ed. Van Impe), 1998, pp. 171-177[7] Thasnanipan, N., Aye, Z. Z. and Boonyarak, T.
“Behavior of Polymer-based Slurry for Deepseated Bored Piles in Multi-layered Soil of
Bangkok”, Proceedings of the 4th International
Geotechnical Seminar on Deep Foundations on
Bored and Auger Piles. Ghent, 2003, Belgium[8] Boonyarak, T., Prongmanee, N., Boonchaiyo,
C., Aye, Z. Z. and Thasnanipan, N. The
assessment of design parameters of wetprocessed bored pile in Bangkok subsoil. The
20th National Convention on Civil Engineering,
Chonburi, Thailand. 2015, ID 348 (in Thai).
[9] Submaneewong C. (1999). Behavior of
instrumented barrette and bored piles in
Bangkok subsoils, (in Thai), M.Eng. Thesis,
Department of Civil Engineering, Faculty ofEngineering
Chulalongkorn
University,
Bangkok, Thailand.
[10] ASTM D 1143, "Standard test method for piles
under static axial compressive load", 1994;
West Conshohoken, P.A.
[11] Aye, Z. Z., Boonyarak, T., Maung, A. W.,
Thasnanipan, N. and Prongmanee, N. “Backanalysis of design parameters for bored pile in
Yangon”. International Conference in SoftGround Engineering, Singapore (ICSGE), 2015,
pp 563-570[12] Schanz, T., Vermeer, P.A. and Bonnier, P.G.
“Formulation and verification of the Hardening
soil model”, in Beyond 2000 in computational
Geotechnics, A.A. Balkema, Roterdam,
Netherlands, 1999, pp. 281-290[13] Benz, T. “Small-strain Stiffness of Soils and Its
Numerical Consequences”. Volume 55 ofInstitut für Geotechnik Stuttgart, Mitteilung.
Inst. für Geotechnik, 2007
[14]Surarak, C., Likitlersuang, S., Wanatowski, D.,
Balasubramaniam, A., Oh, E., Guan, H.,.
“Stiffness and strength parameters for hardening
soil model of soft and stiff Bangkok Clays”.
Soils and Foundations 52 (4), 2012: 682–697.
[15] Likitlersuang, S., Teachavorasinskun, S.,
Surarak, C., Oh, E. and Balasubramaniam, A.
“SmallStrain Stiffness
andStiffness
Degradation Curve of Bangkok Clays”. Soils
and Foundations. 53(4), 2013: 498–509

REFERENCES[1] Tomlinson M.J.,"Pile foundation, The carrying
capacity of piles and pile groups", Foundation
design & Construction 6th edition, 1995, pp.
265-316.
[2] Bowles. J. E., "Pile settlement ", Foundation
analysis and design 5th ed., 1996, pp. 907-909.
[3] Vesic, A. S., "Design of Pile Foundations",
National Cooperative Highway Research
Program, Synthesis of Practice No. 42,
Transportation Research Board, Washington
DC, 1977, 68 pp.
[4] Elaziz, A. Y. A. and Naggar, M. H. E.,

366

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INFLUENCE OF NONWOVEN GEOTEXTILE ON THE HYDRAULICRESPONSE OF MECHANICAL STABILIZED EARTH WALL
Duc Bui Van1, Avirut Chinkulkijniwat2, Suksun Horpibulsuk3, Arul Arulrajah4, Somjai Yubonchit5,
Artit Udomchai6, Irin Limrat7, Tulasi Ram Bhattarai8
1School of Civil Engineering, Suranaree University of Technology, Thailand. Email: buivanduc@humg.edu.vn
Center of Excellence in Civil Engineering, School of Civil Engineering, Suranaree University of Technology,
Thailand;4
Swinburne University of Technology, Melbourne, Australia
5,6,7
School of Civil Engineering, Suranaree University of Technology, Thailand;
8
Senior Divisional Engineer at Ministry of Irrigation, Government of Nepal.
2,3

ABSTRACTThe use of improper drainage system may cause the instability, even failure of Mechanical Stabilized Earth (MSE)
walls. Due to the shortcoming of conventional methods, the drainage system made of geocomposite has been
widely employed as an alternative mean once dealing with the seepage problem. Some of previous works reported
that the water retention characteristic (WRC) of geotextile is similar to that of the coarse-grained soil, however,
the influence of WRC of geotextile on the flow behavior of MSE wall is still not clearly understood. In this research,
the laboratory and numerical investigation was chosen to examine the effect of nonwoven geotextile on thehydraulic response of MSE. Firstly, a large scale physical model of MSE wall was conducted under two conditions
including with and without installation of geocomposite layer. Next, a series of numerical calculation was
performed by using finite element method so as to examine the effect of some soil water characteristics of nonwoven geotextile on seepage responds including the effective saturation, and phreatic surface in the MSE wall
reinforced by geogrids. Results from experiment shows that there is a significant reduction in water level inreinforced zone once the geocomposite employed hence higher stability of MSE obtained. Similarly, parametricstudy is also found out that the distribution of effective saturation at the interface between backfill andgeocomposite material is influenced by the ga parameter whereas the gn plays a little role on the distribution ofeffective saturation at the interface. In conclusion, the profile of effective saturation is mainly affected by the WRCof geotextile, whereas the impact of van Genuchten parameters of geotextile on the phreatic surface is foundinsignificantly. The second factor that affects the distribution of effective saturation along the interface between
soil and geocomposite was observed is that “capillary barrier”.
Keywords: Mechanical Stabilized Earth wall, Geocomposite Drain, Unsaturated Flow, Soil Water Retention.
many of studies [1], [3], [4] indicate that thegeotextiles has WRC similar to those of coarsegrained soils. Currently, various methods have been
introduced to estimate WRC of geotextile including ahanging column test; a capillary rise test approach; anoutflow capillary pressure cell, and modified outflow
capillary pressure cell [5], [6]. In fact, WRC ofgeotextile is now more common; however, the designof geocomposite drains in MSE wall does notincorporate WRC of geotextiles. This paper aims toinvestigate the influence of some soil waterparameters of non-woven geotextile, a large scale ofMSE wall with L-shape geocomposite drains was
employed in this research. A set of instruments wereused to observe flow and deformation responses
during tests. Lastly, parametric study was chosen to
carry out the numerical experiments. The outcomes ofthis research will lead to a better understanding of
which parameters affect the performance of the
geocomposite. This finding can contribute to theeffectiveness of the selection of drainage material,

1. INTRODUCTION
The use of improper drainage system may cause
the instability even failure of Mechanical Stabilized
Earth (MSE) walls. Seventeen of 26 collected casehistories of failure of MSE wall associated to low
permeability of backfill material was investigated by
Koerner, R.M., et al (2001) [1]. After carrying out an
investigation of the disastrous failure of a reinforced
earth wall Yabu city in 2004 Shibuya, S., et al. [2]concluded that the conventional drainage system is
not applicable for high level of ground water.
Due to the shortcomings of the conventional
drainage method such as cost expensive, requires
large site preparation, the use of an artificial drainage
material consisting of a core element having large
flow channel (geonet) sandwiched by nonwoven
geotextile is necessary to provide a drainage system[1]. Despite having many reports on the use of
geocomposite as a drainage system, limited numericalanalysis of MSE walls using geocomposites has beenstated. Similarly, the effect of WRC of non-woven
geotextile has not been taken into account yet, even367

GEOMATE- Bangkok, Nov. 14-16, 2016100

backfill material as well once dealing with seepage
water through the mechanically stabilized earth wall.
2. PHYSICAL AND NUMERICAL OUTLINES
2.1 Theoretical Back GroundOnce
consideringatwodimensional
homogeneous anisotropic material, the transientwater flow within an unsaturated porous medium isgoverned by the following equation
∂θ
∂ 2h
∂ 2h
= kx 2 + k y 2
(1)
∂t
∂x
∂y

Percent finer, %

80
60
40
20
0
10

Where θ is volumetric water content, h is the total
head, kx and ky are the unsaturated coefficient of
permeability in the x- and y- directions, and t is time.
To supplement Eq. 1, constitutive equations related toθ, kx, and ky to h are required. Iryo, T., (2003) [4], andKnight, M., (2001) [5] concluded that there isconsiderable evidence to suggest that van Geuchten(VG) [7] and van Genuchten-Mualem (VGM) [8]models are applicable to nonwoven geotextiles. Thus,
they are employed to approximate WRC andpermeability functions for both the soil and thenonwoven geotextile.
gn gc
S − S resSe === 1 + g a h p(2a)S sat − S res

1
10-1
10-2
Particle diameter, mm (log scale)(a)
Experimental data
Best fit curve
ga =20 [1/m]gn=2.5[-]1000
Suction, kPa (log scale)[ (Poorly graded sand-(SP)Cu=3.08 Cc=0.82

) ]

100
10
1
0.1

k r ( S e ) = S e0.5 [1 − (1 − S e−1 / g c ) − g c ] 2

0.01
0

(2b)
Where Se is effective degree of saturation, S is degreeof saturation, Sres is residual saturation, Ssat is thesaturation of saturated soil, hp is matric suction head,
kr is the relative permeability coefficient. ga [m-1] andgn are fitting parameters, and according to Mualem
hypothesis [9], gc is assigned as 1/gn -1.

20

60
40
Degree of saturation (%)

80

100

(b)
Fig 1. Grain-size distribution (a), and WRC of the soil
used in this research (b).
Table 1. van Genuchten parameters of materials
Notations

2.2 Materialsgagn

The soil used in this investigation is a clean sandwhich consists of 10% gravel, 87.3% sand, and 2.7%
silt. The particle size distribution is presented in the
Fig 1 (a). Optimum water content of 5.7 % andmaximum dry unit weight, γdmax of 16.7 kN/m3. The
saturated hydraulic conductivity of the soil is k= 17
m/day. The WRC of the sandy soil was obtained bymeasuring change of drained water volume during
applied matric suction [9], or capillary force usingpressure apparatus [6].
The hydraulic properties of the geotextile are asshown in Table 1. The transmissivity of the geonet is0.004 m2/sec hence the corresponding saturated
permeability of 69120 m/day. The WRC of the sandy
soil and the geotextile are presented in Fig 1 (b). The
fitting parameters are summarized in Table 1. Iryo, T.,(2003) [4] gathered test results of 13 geotextiles andreported the range of VG and VGM model parametersof geotextiles subjected to both wetting and drying
cycles. Based on [4], the typical range of ga and gn is10-300 [m-1] and 2-6 [-], respectively.

Unit

Soil

Geotextile

Geonet

-1

20
1.5

20
2.5

600
40

[m ]
[-]

Sres
Ssat

[-]
0.03
0.03
0.00
[-]
1.00
0.80
1.00
Geotextile: indicative hydraulic properties
Klateral[m/day] 17
320
69120
Klong

[m/day]

17

2000

69120

2.3 Full-scale test experiments
A large scale physical experiments wereconducted with the dimensions of the tank are shownin (Fig 2). Two physical experiments were employed:
without (case I), and without geocomposite drain
installation (case II). After filling sandy soil into thephysical model up to height of 1.4m, the upstream
water was increased from the initial level to surfaceof 0.0m, 0.8m. 1.1m, 1.5m, respectively. An acrylic
plate was employed to make the wall facing. 5 layersof bearing reinforcements [10] of equal length of 0.7
m [11] were installed in the reinforced zone.
Locations of the instruments are illustrated in Fig 2.

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GEOMATE- Bangkok, Nov. 14-16, 2016

(a)

(b)

Fig 2. Sketch of the physical model and location of the instruments (a) Plan view and (b) Side view
Table 2. VG and VGM model parameters assigned to every cases in the numerical experiment
Sres
(-)

0.03

0.03

Ssat
(-)

1.0

1.0

gn
(-)

1.5

1.5

Sandy soilgakx(m-1) (m/day)

20

20

17

17

ky
(m/day)

17

17

Sres
(-)

0.03

0.03

Ssat
(-)

0.8

0.8

gn
(-)
2.2
2.5
3.0
4.0

2.5

Geotextilegakx(m-1) (m/day)

20

16
20
40
100

320

320

369

ky
(m/day)

2000

2000

Sres
(-)

0

0

Ssat
(-)

1.0

1.0

gn
(-)

40

40

Geonetga
kx(m-1)(m/day)

600

600

6.9E4

6.9E4

ky
(m/day)

6.9E4

6.9E4

Remarks
Case 1Case 2
Case 3
Case 4
Case 5
Case 6
Case 7
Case 8GEOMATE- Bangkok, Nov. 14-16, 2016

(a)

(b)
Fig 3. Phreatic surface for MSE wall (a) Case I and (b) Case II370

GEOMATE- Bangkok, Nov. 14-16, 2016

(a)

(b)
Fig 4. Mesh discretization of the models (a) without, and (b) with- geocomposite
with depth for all ga values and are found to reachfull saturation at a depth of 1.0 m. It is also clearlyshown that a greater ga exhibits a greater degree ofsaturation along the interfacial section. The
saturation profiles at shallow depth are found to beslightly different from each other, but it then notably
increase at deeper depth. At the interface between
the soil and the geotextile, the capillary barrier will
occur if the water flows from a low ga to high ga
materials [12], and hence the accumulation of watertakes place along the interfacial section. Thisphenomena “capillary barrier” occur until thepermeability of the two materials reaches the samevalue or the magnitude of suction at the interfaces
reduces to a critical suction and water breaks throughthe interface [12].

2.4 Numerical Experiment Outline
The finite element mesh, boundary condition are
shown in Fig 4(a) and 4(b) for the MSE wall without
and with geocomposite drain installation,
respectively. 15-node triangles were assigned to themodels. The parameters for the hydraulicconstitutive equations VG and VGM for the sandy
soil, the geotextile and the geonet assigned to themodels are summarized as shown in Table 2. In total,
08 cases of numerical simulation were conducted inthis study using Plaxis 2D, 2011.
3. RESULTS AND DISCUSSION
3.1 Flow Through MSE Wall Results
Fig 3a, and 3b show the water level at different
upstream water levels for the tests without and withgeocomposite installation, respectively. At any
height of upstream water level, the water level
decreases through the wall face. The significant
reduction in water level as well as water table in theprotection zone shown in Fig 3a, 3b may result inhigher stability of the wall.
3.2 Effect of water retention character of
nonwoven geotextile
3.2.1 The van Genuchten parameter ga
Figure 5 presents the effective saturation profiles
along the interface between the soil and thegeotextile for various magnitudes of ga of thegeotextile. The effective saturation profiles increases

Fig 5. Effective saturation profiles along theinterface for various magnitudes of ga of geotextile.
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GEOMATE- Bangkok, Nov. 14-16, 2016

Critical suction, kPa

Fig 6 presents the critical suction at different valueof ga of geotextile. The magnitudes of the criticalsuction for all cases are quite so low thet water is notable to permeate through the interface at high suctionlevels. Thus, the saturation profiles at high suctionlevels are not so different for all cases (Fig. 5). A
greater ga of the geotextile yields a lower magnitudeof critical suction, subsequently results in a higher
saturation at high saturation levels.
5“capillary barrier” can take place once water flowfrom a low value of ga to higher one, this phenomena
plays a vital role in the distribution of effective
saturation at the soil-geotextile interface. The lower
magnitude of suction where the water permeates
through the interface results in a greater amount ofwater accumulation at this interface.
3) Inversely, the parameter gn does not affect thehence the distribution of effective saturation.
ACKNOWLEDGEMENTS
This work was financially supported by Thailand
Research Fund under MRG, the TRF SeniorResearch Scholar program Grant No. RTA5980005,
and Thailand Toray Science Foundation, and the
SUT-Ph.D scholarship Program for Asean.
REFERENCES[1] Koerner, R.M., 2001. “Geosynthetic Reinforced
Segmental Retaining Walls”. Geotextiles andGeomembranes 19, 359-386.
[2] Shibuya, S.,et al., 2007. “Failure of Reinforced
Earth as Attacked by Typhoon No. 23 in 2004”.
Soils and Foundations 47, 153-160.
[3] Bouazza, A., et al., 2006. “Water Retention of
Nonwoven Polyester Geotextiles”. Polymer
Testing 25, 1038-1043.
[4] Iryo, T., 2003. “On the Hydraulic Behavior of
Unsaturated
Nonwoven
Geotextiles”.
Geotextiles and Geomembranes 21, 381-404.
[5] Knight, M., Kotha, S., 2001. “Measurement of
Geotextile-Water Characteristic Curves Using aControlled Outflow Capillary Pressure Cell”.
Geosynthetics International 8, 271-282.
[6] Nahlawi, H., et al., 2007. “Characterisation of
Geotextiles Water Retention using a Modified
Capillary Pressure Cell”. Geotextiles andGeomembranes 25, 186-193.
[7] van-Genuchten, 1980. “A closed-form Equation
for Predicting the Hydraulic Conductivity of
Unsaturated Soils”. Soil science society ofAmerica journal 44, 892-898.
[8] Mualem, Y., 1976. “A new Model for Predicting
the Hydraulic Conductivity of Unsaturated
Porous Media”. Water Resour. Res 12, 513-522.
[9] Handoko L, et al. 2013. “Suction controlled
triaxial apparatus for saturated-unsaturated soiltest”. International journal of GEOMATE:
geotechnique, construction materials andenvironment 4(1):466-470.
[10] Horpibulsuk S. 2011. “Performance of an EarthWall Stabilized with Bearing Reinforcements”.
Geotextiles and Geomembranes 29(5):514-524.
[11] AASHTO, 2002. “Standard Specifications for
Highway and Bridge”..
[12] Ross, B., 1990. “The diversion capacity ofcapillary barriers”. Water Resources Research
26, 2625-2629.

4
3
2
1
0

20

40
60
80
100
van Genuchten parameter ga

Fig 6. Relationship between critical suction and ga
3.2.2. The van Genuchten parameter gn
Figure 7 presents the effective saturation profiles
along the soil-geotextile interface for various
magnitudes of gn. It is an evident that the gn
parameter plays a little role on the effective
saturation profile along the soil-geotextile interface.
The parameter gn does not affect the pore size ofmaterial, hence capillary barrier, resulting in a little
difference in the degree of saturation at the interface
between soil and geocomposite system.

Fig 7. Effective saturation profiles for various
magnitudes of gn of geotextile.
4. CONCLUSION
The
experimental
results
revealthat
geocomposite can be employed to prevent waterflows into the reinforced zone effectively. Similarly,
leading to the following conclusions:1) The effective saturation profile in the MSE issignificantly affected by WRC of geotextile.
2) At the interface between two materials having
different pore size, a phenomena known as372

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STABILIZING A STEEP WEATHERED SANDSTONE SLOPE ALONG
THE RAILROAD IN SOUTHERN GERMANY
Thomas Hangartner1 and Christophe Balg2
Geobrugg AG, Geohazard Solutions, Switzerland; 2 Geobrugg AG, Geohazard Solutions, Switzerland

1

ABSTRACTDue to the fact that the German Railroad reduced the maintenance work on the important railway line
Nuremberg – Regensburg, section Deining - Batzhausen they had several cases with blocks laying at the tracks.
Therefore they decided to stabilize the 70° steep weathering prone sandstone slope. As a protective measure theflexible slope stabilization system using high-tensile steel wire meshes was selected in combination with nailing.
This widely used way to stabilize soil and rock slopes is economical and a good alternative to shotcrete solutions
or massive supporting structures. Special concepts have been developed for dimensioning of flexible surface
stabilization systems in steeper soil or heavily weathered loosened rock slopes, but also on jointed and layeredrock in which the bodies liable to break out are determined by joint and layer planes. Analyzing of the design,
explanation of the installation and how the finished installation behaves will be shown. The successful
implementation confirmed that flexible slope stabilization systems by the means of high tensile steel wire
meshes is well suitable for practical applications. The system proofed to be fast installed, adaptable, aesthetic
and economic.
Keywords: RUVOLUM, TECCO, dimensioning, flexible slope stabilization system

INTRODUCTION
PROJECTThe use of flexible slope stabilization systems
have proven there suitability around the world,
including Europe, Asia, North America and in colder
climates, where the stabilizing facings need to beable to flex under the freeze/thaw cycle. Historically,
the mesh used for these purposes is produced using
mild steel wire with a tensile strength of 400–500
N/mm2. The development of mesh made from hightensile steel wire with a tensile strength of at least
1770 N/mm2, offers new possibilities for theefficient and economical stabilization of slopes (seeFigure 1). Sophisticated dimensioning concepts
serve to dimension these kinds of slope stabilization
systems against superficial instabilities by taking thestatics of soil and rock into account [1].

On the important German railway line
Nuremberg - Regensburg, section Deining Batzhausen they had several cases with blocks
laying at the tracks afterwards they reduced themaintenance work on this section. Therefore theydecided to active stabilize the 70° steep, up to 15 mhigh and weathering prone sandstone slope. A
protective measure had to be selected to stabilize the8500 m2 of the exposed cutting against superficial
instabilities, tilting as well as sliding of individual
blocks and rockfall.

Fig. 1High-tensile wire mesh for slope
stabilization and system spike plate to tension themesh against the slope surface.
373

GEOMATE- Bangkok, Nov. 14-16, 2016

TECCO G65/3 mesh has a tensile strength of
150 kN/m,which means substantially higher forces can beabsorbed by this mesh in comparison toconventional mild steel wire mesh. Aside from thehigher bearing capacity, another advantage of
TECCO mesh over conventional mild steel wire
mesh is that it has an even load transmission and noweak zones within the mesh. This is achieved bymanufacturing TECCO mesh with the same
diameter high tensile wire, which forms a unified
and homogenous mesh structure [4].

Fig. 2
Location of project in the area of
Deining in southern Germany.

Special diamond-shaped system spike plates
which match the load capacity of the mesh serve to
fix the mesh to soil or rock nails. By tensioning
these nails, and recessing the spike plates into theground, the mesh is adequately tensioned to ensure it
follows the surface contours [4].

Fig. 3
Partly eroded rock slope before theinstallation works.

With this slope stabilization system the rows ofnails are offset to each other by half a horizontal naildistance. This limits the maximum possible breakout between the individual nails to a width “a” and alength of “2 x b” (see Figure 5a). The staggeredlayout is shown in Figure 5b for the project
Dongcheon in Korea.

Table 1 Parties involved
Client:Project:
Nailing andsysteminstallation:
Date ofinstallation:

German Railway AG, Nuremberg,
Germany
CDM Consult GmbH, Munich,Germany
SPESA GmbH, Schrobenhausen,
Germany
July 2009 – May 2010 (including
winter break)Fig. 5aFig. 4
Typicalprotection measure.

cross

sectionGeneral profile with nail arrangement [4].

including

PROTECTION MEASURE
The flexible slope stabilization system consistsof TECCO G65/3 high-tensile steel wire mesh,system spike plates and soil nails. The mesh is madefrom 3 mm high tensile wire and uses a zincaluminum coating for protection against corrosion.
Each diamond of the single twist mesh measures 83
mm x 143 mm. The high tensile steel wire used in
the manufacture of the mesh has a tensile strength of1770 N/mm2, compared to mild steel which has atensile strength of 400–500 N/mm2. As a result

Fig. 5b Staggered pattern of nail installation, project
Dongcheon, Korea.
374

GEOMATE- Bangkok, Nov. 14-16, 2016

DIMENSIONING
The flexible slope stabilization system wasdimensioned against superficial instabilities basedon the RUVOLUM concept [2, 3]. The maximum
nail spacing and the required nail length can be
determined, and by utilizing the high bearing
capacity of the mesh, significant cost savings can berealized by reducing the number of nails required.
Conventional slope design methods are still requiredfor deeper seated failure mechanisms.

Fig. 7 Scaffold for drilling after cleaning of
the slope.

Fig. 6 The dimensioning concept is based on theinvestigation of superficial slope parallel instabilities(left) and on the investigation of local instabilities
between single nails (right) [2, 3, 7].
Table 2 Project information
Height of the slope:
Subsoil:

Inclination of theslope:
Stabilized area:
Nail type:

10 - 15 m
Variable weathered
sandStone (partly eroded)
70°

Fig. 8 Beginning of the mesh installation.

8’500 m2
Gewi ø32 mm

INSTALLATIONFirstly, the slope was cleaned of eroded soil andsmaller loose rocks. Due to the fact that there was
no access to the top of the slope, the installationcompany decided to install the nails fromscaffolding (see Figure 7). It was very importantthat the nails could been installed in deep seatedspots so that the mesh could been tensioned and kept
in contact with the surface. Shotcrete was used at
locations where undercutting was occurring, tofurther secure any blocks from sliding down.

Fig. 9 Overview and nail pattern adapted to
the local situation.

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GEOMATE- Bangkok, Nov. 14-16, 2016

When slopes stabilized with flexible high tensile
steel wire mesh are combined with erosion control
mats, they can regain a natural or vegetatedappearance, which aesthetically is normallypreferred.
Depending on the country, the price of flexible
slope stabilization systems is equal or lower incomparison to reinforced shotcrete solutions
including similar fully mortared rock bolts of thesame length and diameters.
Fig. 10REFERENCES

Adaptability of the installed mesh.

REVEGETATION/ EROSIONPROTECTION[1] Brändlein P. (2004). LGA Nuremberg,

Erosion control mats can be installed underneaththe mesh to aid in revegetation. The application of avegetation layer can be limited by the soil or rockproperties, groundwater and climate. The steeper
the slope cutting, the more difficult it becomes forvegetation to grow. If revegetation is to be carriedout, a species of plant or grass should be selectedthat is fast growing and suitable for the local
conditions [4, 6].

Germany, Monitoring and supervision oflaboratory testing of the TECCO slope
stabilization system, Test report BPI 0400046/1.
[2] Rüegger, R.; Flum, D. (2006). Anforderungenan flexible Böschungsstabilisierungssysteme
bei der Anwendung in Boden und Fels.
Technische Akademie Esslingen, Beitrag für 4.
Kolloquium „Bauen in Boden und Fels”.
[3] Rüegger, R.; Flum, D.; Haller, B. (2002).
Hochfeste Geflechte aus Stahldraht für die
Oberflächensicherung in Kombination mit
Vernagelungen und Verankerungen.
Technische Akademie Esslingen, Beitrag für 2.
Kolloquium „Bauen in Boden und Fels“.
[4] Cala , M.; et al. (2012). TECCO® Slope
Stabilization System and RUVOLUM®
Dimensioning method. AGH University of
Science and Technology.

Fig. 11[5] Brisbé York, A.; et al. (2006). RUVOLUM

Condition of the slope in July

Dimensioning Software 7.0, Spain and2011 one year after successful installation.

Switzerland.
[6] Mrozik, M.; et al. (2006). Stabilization of a 30m deep cutting along the highway A63
Kaiserslauten – Mainz, Germany, with aflexible slope stabilization system consisting ofhigh tensile steel wire mesh in combination
with nailing. Autostrada Polska.

CONCLUSIONThe TECCO slope stabilization system can beadapted to the site specifics and static conditions in avery flexible manner. The system can be designed
and dimensioned against superficial instabilities,
which is the first time flexible surface support
measures can be properly designed. This approach
offers the possibility to arrange the nails in a more
economical way due to the capability of TECCO
mesh in absorbing and transferring high loads.

376

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

REFLECTION OF TECHNOLOGIES OF NAVIGATION IN SPACETIME IN THE STRUCTURE OF ARCHEOLOGICAL OBJECTS
1

Alina Paranina1
Faculty of geography, Herzen State Pedagogical University, Russia

ABSTRACTThe article shows the objective criteria of sacralization of objects of natural and cultural heritage of Eastern
Europe and Southern Siberia, based on the performance of informational function in ancient human life-support
system (as instruments of navigation in space-time). Transformation of functions of sacral objects from their
inception to the beginning of the third millennium, due to the development of geo-cultural space (changes intechnology, socio-cultural paradigm and social and economic conditions). The stages of the evolution ofnavigation technologies, selected by the author on the basis of cultural layering of archaeological objects arereviewed in the article. The carried-out retrospective analysis allows to make the assumption that improvementof technologies of orientation could be a basis of emergence and development of Homo sapiens (sapientation):astronomical supervision in horizon observatory develops a system of ecological thinking; supervision over a
shadow of the gnomon tool develops abstract thinking (the abstract graphic sign is genetically connected withconcrete natural process)Keywords: Navigation, Adaptation, Modeling, Technology, Evolution.

INTRODUCTIONOBJECTS AND METHODS OF RESEARCH

The navigation concept of information modelingof the world, developed by the author, explains theancient sacred status of natural and cultural heritageby their important role in addressing the adaptationof the geographical challenges – namely in thedevelopmentofgeographicalspace-time.
Application of methods of cultural geography andpaleogeography,
paleoastronomy
andastroarcheology allows us to prove that these objects
were created as orientation tools in space-time andprovides the opportunity to "be at the right time in
the right place," i.e. performed the informational
function in life-support system [1]–[7].
Interdisciplinary studies of natural and cultural
heritage of the Eastern Europe and Southern Siberia,
conducted by the author in 2009-2015, together witha specialist in the field of archeology, culture,
astronomy, showed that: 1. high degree ofcorrelation of the dominant trends of the spatial
structure of the ancient sacred objects with calendar
azimuths of sunrise / sunset (and other astronomicalreference points); 2. coincidence of these areas withthe planned figure of the landscape; 3. reflection in
the organization of sacred space (in the size ofobjects and the distances between them) of thesystem of ancient metrological standards.
The tradition of inheritance underlines theabsolute importance of the position in geographic
space, its function in the intergenerational transfer
of knowledge – "comtinuity", and allows us to trace
the development of navigation technologies.

For the study were chosen representativearchaeological sites (most typical and well studied
by standard methods of the natural sciences and thehumanities), revered to date: stone labyrinths of theWhite Sea, revered stones in the North-West Russia,
the archaeological complex Divnogorie (Voronezh
region), petroglyphs and megalithic complexes
Gorny Altai, sacral complex Salbyksky valley(Khakassia), and others [8]–[12].
Field work, cartographic materials, aerial photos,
satellite imagery and Google materials, scientificpublications for the projects of natural and cultural
heritage and the research problem were used toanalyze the objects.
The applied methods of field of geographical(topographic,
geologicalandstudiesgeomorphological landscape and geomorphologicalsurvey), working with thematic maps and remotesensing data (landscape interpretation, selection anddescription of lineaments, planetary fracture
systems, directions of outline structure of artificial
and natural facilities with the construction ofdiagrams of roses), standard laboratory analyticaland statistical methods, the methods of themetrological analysis of archaeological objects andastronomical
paleoastronomical
calculations(altitude and azimuth position of the sun – with thehelp of astrocalculator, length and direction of the
shadows – using basic trigonometric functions), as
well as methods of mathematical, cartographic andconceptual modeling were used.
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GEOMATE- Bangkok, Nov. 14-16, 2016

EVOLUTION OF TECHNOLOGIES
Analysis of Structure of Objects
Stability and continuity of the information
function of natural and cultural heritage are provedby numerous examples of the combination of objectsof different ages and modern geodetic network: in
the North and North-West Russia – "landscapemarkers and halt points - stone labyrinths - geodetic
points"; in Khakassia in "outrigger stele mound(ancient) - mound (later time) -geodesic point" andothers. In spatial structure of the studied ancient
objects azimuths by which dates of equinoxes andsolstices are determined are revealed (Fig. 1-6).

Fig. 1 The plan of
Salbykskaya valleys.

asacralcomplex

Fig. 3 Megalytic complex of the esteemed stone
the St. Paraskev (Leningrad Region, lake Vrevo).

Fig. 4

Plan of a complex St. Paraskev.

Fig. 5

Stone labyrinth- gnomon of the White Sea.

in

Ряд1
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Fig. 6 The schedule of a shadow of a gnomon in ayear. Row 1-12 – schedules of every month.

Fig. 2 Comparison of the azimuths measured
between barrows and in a surrounding landscape.
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GEOMATE- Bangkok, Nov. 14-16, 2016

Main Stages
Technologies

ofEvolutionofare easily determined azimuths of linear "chains" ofstone objects, providing the definition of theastronomicalseasonsboundaries.
Obviouspossibilities of astronomical use of the natural andnatural-artificial sculpture, based on their position in
space, are an additional confirmation in the
traditions and legends.
Determining the proportion of creative humanparticipation in the formation of animal andanthropomorphic rock sculpture is simple enough: acomparison with neighboring geological andgeomorphological complexes of similar genesis
shows that the principle of the ancient architects was
to "remove the excess." Apparently, this principle
developed from the ancient taboo on the number ofstrokes taken on the stone to the quest for a perfectmatch to the living prototype (in classical sculpture).
The wisdom of ancient technology, reflecting thementality of "co-creation with nature", limiting thedegree of change in the natural substrate, highlights
the problems of modern interaction between humanand nature as a bold violation of the natural forms of
the earth's surface leads to a change, sometimes
catastrophic, of water and air flow – those naturalprocesses that remain the major sculptors of theplanet.
The later artificially established megalithic sites
can be considered as one of the stages ofdevelopment of this tradition.

Navigation

Based on the analysis of different age elementsof sacral complexes, objectively there are five stages
in the development of navigation technology:
Landscape and megalithic (direct sight), megalithic(back sight), the historic stage of abstract modeling(development of a rational and creation of irrational
symbolic), the current stage of the new navigation
technology, communication and modeling.
Landscape stage of development of space-timeLandscape stage of development of space-time,the initial – describes the utilization of geographicalspace for fixing astronomical directions - space-timebenchmarks. Foresight (observation), then check thesunrise / sunset at the solstices and equinoxes (andother astronomical targets) contribute to: 1. open and"rugged" horizon heterogeneity which allows us torecord significant azimuths 2. lineaments – linear
elements of tectonic structures, topography, linear
borders of aquatic and territorial natural complexes.
At this stage of development of orientation
technology the landscape serves as astronomical
instrument, and to keep the life-saving information
the energy of natural processes, which indefinitely
long maintain the spatial structure, is used.
To fulfill the information function of thelandscape, since ancient times, with the economic
zoning of the territory they received a special sacredstatus, the main feature of which is the minimum
load in order to maintain the stability of the spatial
structure. Thus, the geographical criterion ofsacralization of the landscape is a role in the life
support system, which is determined by the need forhuman resources, information on the procedure ofspace-time. This finding is consistent with the
ethnographic data on the exclusive role of therevered natural objects in the pre-Christiantraditions, dedicated to the Sun and the calendar.
The traditions of minimal impact on sacredlandscape in the form of its labeling with stable tothe destruction of cells or artificial additions(petroglyphs, petroglyphs on the rocks, the additionof rough stone –“obo”, symbolic gifts in the form ofribbons or trophies fragments) are preserved. Shapeand density of man-made additions into the sacredlandscape orientation reflect the development oftechnology and especially the mentality, regional
specificities, in particular, the availability of humanactivities in energy, raw materials and information
resources, the demand for navigation information(social order).
Often, natural formations – erosional forms andremnants of weathering, associated with various
forms of living beings. The issue of their natural or
artificial origin is discussed, but on satellite images

The second stage of development of navigation –megalithic
The second stage of development of navigationand information functions of the location of thesacred complex is megalithic, characterized by theuse of artificial tools for foresight, fixing the hairlines of main astronomical azimuths and theirrelationship with the iconic elements of thelandscape and visible objects on the horizon by aresistant to destruction and displacement of objects .
The requirements of this technology meet the largemegaliths objects placed at a distance of a few
hundred meters. Location of the horizon observatory
is possible both on high, visually-related points of
the relief, and in the valley – decrease, nearing thehorizon, which makes landfill compact andrelatively independent from the surrounding
landscape changes. Exemplary embodiments of thistechnology can serve megalithical complexes of the
Caucasus, the Kola Bay (megaliths city park plateau
Sade and Crow stone in Murmansk) and others.
Even if the material of the megalithic complex
was not subjected to significant relocations andprocessing, it should be seen as artificial, naturaland man-made, because the process of creation ofastronomical tools embedded knowledge and humanlabor. Different size and degree of treatment may
indicate the value. One of nearly invisible, but
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GEOMATE- Bangkok, Nov. 14-16, 2016conclusive evidence of artificial stone, or selection
criteria, are measures, multiple units of the ancientsystem of measures – feet, elbows, fathoms.
Metrological analysis is needed to define thedistance between objects.
In Salbyksky Valley megalithic stage left behindthe unique objects ("statuesque", complex "twostones", stone of fertility in the gate of the Grand
Salbyksky mound), and the same type outrigger stele
located on the west side near the largest burial
mounds. Megalithic sites are the oldest in the valley,
on the size, shape, quality and extent of thedestruction of the stone, they are very different fromthe material used at a later time to build mounds.
Similarly, in other regions – megalithic sites are themost ruined and slightly different from the naturalelements of the landscape. So, seids are often
perceived as typical "glacial boulders" and tounderstand "the structure due to function", acomprehensive eco-geographical and astronomical
analysis is required, and for the definition of featuresof "architectural style" – a video, including local andregional variants (e.g., seids of Kola Peninsula,
Siberia, North and South America).
At the landscape stage secure fit of astronomical
directions was ensured by use of natural rock masses– the most resistant to tectonic movements and shifts
under the influence of exogenous geomorphologicalprocesses (e.g., slope or cryogenic), at the megalithic
stage, the maximum non-mobility of artificiallyestablished monolithic sights was provided by theirweight, as weight determines the dynamics ofobjects. The gigantic sizes of buildings such aspyramids or temple complex of Baalbek wereconsistent with this task.
During the first two stages of astronomical
observation people could see that horizon
observatory after time unavoidably "lose theiraccuracy," not because of their design or underlying
rocks are unstable, but due to the fact that theastronomical azimuths of celestial landmarks
climaxes gradually changed, i.e., Space itself is
dynamic. On the basis of accumulated knowledge it
has become possible to calculate the long-period
cycles and switching to more efficient, compact andprecise technology, where the temple-observatory is not just a tool, but it is a fixed point in space, theplatform and data warehouse (calendar layout andmarkings).

which the shadow of one object to another object
falls at sunrise / sunset (at least – of the Moon) on
certain days of the year. In this case, the totality
creates the functional integrity of forms – local
astronomical and geodetic network. Two objects on
the line W-E create the most simple form a network,
which are linked by shadow twice a year – at theequinox, at sunrise and sunset. This approachrevealed the solar calendar function in relief imagesof labyrinths in Galicia [13].
Observations of the shadow of the subject – thegnomon, not only at the beginning, but also
throughout the light of the time of day, gives aperson a lot more information – measuringhorizontal and vertical angles, accurate clock andcalendar binding to the planetary system of
coordinates (latitude), the transition from local
astronomical and geodetic network technology tocreate regional networks. The value of switching tonavigation technologies of backsight emphasized in
all cosmogonic myths, like the creation of the World
from Chaos. As a rule, the myth explains thedetermination of the center of the world – axis,
navel, etc., in other words, the center of a regional
astronomical and geodetic network, the fixed
position of Omphale.
Inclusion of objects into the regional network
explains their position on the same latitude of othersacred objects. Thus, Divnogorie and Salbyk arelocated at the latitude of 51-52º, close to the latitudeof Arkaim, Kiev and Stonehenge. This latitude
corresponds to one of the boundaries of the sevenclimates, recorded in written sources of Ancient
Babylon. At the latitude of St. Paraskeva there are:stone in the Pskov region, similar in structure andname, Shum-Gora (the largest mound in Europe,
presumably – the tomb of Ryurik), Peryn (theancient sanctuary on Lake Ilmen in the Novgorod
neighborhood). The widespread development oflocal and regional networks provided toponymic
marking of territorial navigation systems in spacetime by the sun – the principles of a navigational
markings developed on the historical geography of
the works of V.I. Paranin [1], [2].
The largest number of heritage sites, performing
the functions of a sundial-calendar, were revealed
for the last 5000 years (labyrinths, stele, idols),
perhaps because the older have not survived or have
not yet been investigated. The oldest archaeologicalfinds are: Shigir idol – 11,000 years, the sundial of
England and Turkey – 10,000 years, and others. The
largest amount of solar labyrinth-calendars isconcentrated in the Northern Europe. Calculations
show that, at the southern orientation of the entrance
to the labyrinth, its stone arcs serve as marks of theheight of the midday sun, in other cases, the calendar
information azimuths fixed entrance and all the
spirals.

The third stage is the backsight technology
Backsight has many embodiments, among which
the most famous is path technology of fixation of
shadow of the object and a focused beam. It is clearthat the transition to monoinstrument – the gnomonof a sundial and the calendar took place gradually,
through a series of intermediate options. For
example, there are still a lot of megalithic sites, in380

GEOMATE- Bangkok, Nov. 14-16, 2016

The biggest and most ancient cluster of stonelabyrinths (Stone Age) is located on the island of
Bolshoy Zayatsky, Solovetsky archipelago, andindividual objects and images – on all continents
except Antarctica. Labyrinths were built, as a rule,on the waterways, settlement sites, near freshwater
sources. Examples of distance from the coast are notfrequent, one of them – the labyrinth of the Bronze
Age in Mostishche, 15 km from Divnogorie. Many
labyrinth images were found on stone slabs inDagestan.
An example of embodiment of a backsighttechnology is just a combination of "stone-tree",where the shadow of a tree falls on a stone at a
certain hour. Tradition of honoring of the sacredstone and wood is preserved until recently in theBaltic regions and in Karelia.
The development of navigation is always
accompanied by the process of creating an iconic
space-time equivalents and rational life-affirming
symbols. Undoubtedly, the foundation of sign
systems is made on the basis of foresight. According
to V.B. Frolov, the basis of the primitive graphics in
the Old Stone Age was fused astronomical andmathematical knowledge. However, the mostproductive source of signs should be considered as
the shadow of the gnomon, broadcastingnavigational information in some regions almost
continuously (in the conditions of the polar day).
Analysis of the semiotics of light / shadow allows toreconstruct the information model of the world,which is different from the mytho-poetic model ofhumanities research that sets a measurable
quantitative relationship between the natural process(lighting mode) and the spatial structure of culturalobjects (layout structures and graphical basis ofsigns and images).
In general, the progress of technology ofnavigation of the first three phases expressed in
improving the quality, while saving money andspace requirements. As a result, vitally important
information about the space-time has been inherited,
compressed in form and developed content.
However, at the beginning of the Iron Age, in theconditions of the artificial urban environment,
development of culture takes on a new humanitarian direction. History tells us that the
specific feature of the urban culture and thedevelopment of state forms of social order becomes
progressive idealism and manipulation ofknowledge, including the calendar systems.

world with the creation of rational – reflecting realnatural processes and irrational characters –reflecting speculative build in the delimitation of thenature of cultural space (even if this limitation – walllaboratories and / or experimental conditions).
Technologies of navigation underlying the lunarsolar calendar, cease to be a priority for scientificand technological direction, falling back to morestable – distant landmarks of the sky of stars. The
realization of the principle of the constructive role ofthe super-system of the universe with respect to thesolar system finds its strain – Idealistic expression.
There comes a stage of oblivion of achievements
and transformation of meanings of the ancient
culture based on solar navigation. Created at thattime, the objects relate mainly to the cultural
heritage category: palaces and religious complexes,
but the navigation continues the tradition of binding
object to create a space-time coordinates using thegnomon of the "laying" stone.
We know with what bitterness dominant religion
eradicated the natural urban culture that hasdeveloped as the experience of hundreds ofthousands of years: the baptism and enslavement ofRus, the Inquisition – In Western Europe. However,
the traditions associated with the life support arevery stable. In Christianity, for example, everythingthat could not be destroyed, was lit and left there
under a new – Christian name. In this case, objects
that once served as tools to ensure the real nature,according to the order of social life in the new
system could save only a formal function - tomaintain the tradition of ancestor worship and allow
symbolic treatment to universals of Time(maintenance of health, prolong life, and the good
childbirth – motives of reverence stone, extant).
About the former connection to sight and eye, it
resembles only water uses from these stone buckets"to improve vision."
Fifth - the current stage of the latest technology
The current stage of the latest technology,
characterized in that the ways of orientation, basedon the achievements of the space, computerequipment, development of remote sensing data andcommunication systems develop actively.
In the era of the dictates of the world market andimposed on them by the state, mass consumer
culture, the origins of modern culture forgotten
objects of natural and cultural heritage of theveneration of objects are transformed into
recreational resources (tourism) and othercommercial activity (a wide range of health andmagical practices).
Today, numerous examples can prove that
almost all the heritage sites are focused and related
to the calendar. However, the success ofastroarchaeological and paleoastronomical studies

Fourth - the historic step of abstract modeling
The historic step of abstract modeling ischaracterizedbyadvancesinnavigation
infrastructure of geo-cultural space, differentiationof features of tools (magnetic compass, water andhourglass), the development of abstract modeling381

GEOMATE- Bangkok, Nov. 14-16, 2016(e.g., decoding of astronomical features ofStonehenge or the pyramids of Egypt) receives
essentially idealistic interpretation – considered in
the context of the local activities of the cultcharacter. This line of thought obviously inherits
practices and ideological direction, formed at thestage of the merger of church and state. It affects thesame lack of scientific methodology, based on thelaws of natural diversity, the weak development ofmethods aimed at identifying cause-and-effectrelationship, the traditional art criticism forhumanities research focus.
A variety of scientific methods used fornavigational objects historic and prehistoric past,
involves interdisciplinary cooperation with a view tothe mutual enrichment of Sciences and the formationof multifaceted conception of the problem of naturaland cultural heritage [1]–[15].
Modern science and engineering technologies
enable the broad scientific research of thenavigational designation of ancient objects.
Geographical information system analysis ofprimary navigation purpose of sacred objects ofnature and culture shows that objects are unique andtheir technologies are versatile, as technologies
reflect the laws of the structure and functioning of
the geographic space-time, and record the individual
objects unique combination of geography at every
point. The collection of iconic qualities of thelandscape, representing the real basis of science andpractice, become an integral part of the content ofsacred geography, semiotics of geographical spaceand geography culture. Interdisciplinary studies ofnatural and cultural heritage in the geography will
ensure the conservation status of objects, the correct
definition of their composition, structure andboundaries, the rational use for tourism andrecreation.

REFERENCES[1] Paranin VI., Historical Geography of thechronicle of Russia. Petrozavodsk: Karelia,
1990.
[2] Paranin VI., The history of the barbarians, St.
Petersburg: RGS, 1998.
[3] Paranina GN., Light in the labyrinth: time,
space, information. St. Petersburg: Asterion,
2010.
[4] Paranina GN., “Northern labirinths – gnomon
and models of geographical space”. J. Procedia
Social and Behavioral Sciences. ELSEVIR, Vol.
19, 2011, pp. 593-601.
[5] Paranina AN., “Northern labyrinth – a key totime, space, information”. J. Natural Resources,
Vol. 4, 2013, pp. 349-356.
[6] Paranina AN., “Navigation in space-time as the
basis
for
Information
modeling”.
J.
Archaeological Discovery, Vol. 2, N3, 2014, pp.
83-89.
[7] Paranina AN., “Gnomon as sours of information
on planet rhythms”. J. Geomate, Vol. 10, 2016,
pp. 1815-1821.
[8] Paranina AN., “Reflection of space-time inancient cultural heritage of Urals”. Open AccessLibrary Journal. Vol. 3, 2016, e2888.
[9] Paranina АN., Paranin RV., “Navigation in
geographical space as a factor of developmentof civilizations”. in Proc. Int. Conf. on ICAE2015 (Applied Ecology: Problems, Innovations)2015, pp. 211-215.
[10] Paranina АN., Paranin RV., “Northern
labyrinths as navigation network elements”
Activities in Navigation. Marine Navigation and
Safety of Sea Transportation. Ed. A. Weintrit.
London-New-York-Leiden: CRC Press, 2015,
ch. 4.
[11] Marsadolov LS., Paranina GN., “Salbykskyarchaeological complex as an object of naturaland cultural heritage”. J. Izvestia RGO, T.143,
Vol. 2, 2011, pp. 79-90.
[12] Marsadolov LS., Paranina GN., “Megaliths as
objects of cultural heritage of the North-West of
Russia”. J. Izvestia RGO, T.143. Vol. 6, 2011,
pp. 57-65.
[13] http://oestrymnio.blogspot.com.es/2012/12/labe
rintos-de-mogor-marin.html[14] Sutriyono E., Hastuti EWD, Susilo BK.,
“Geochemical assessment of late paleogene
synrift source rocks in the South Sumatra
basin”. J. Geomate, Vol.11, 2016, pp. 22082215.
[15] Zaalishvil V., “Spectral characteristics ofseismic waves at strong ground motions”. J.
Geomate, Vol. 10, 2016, pp. 1706-1717.

CONCLUSION
The spatial structure of the sacred (revered)objects of natural and cultural heritage provides
opportunities
for
instrumental
astronomical
orientation in space-time. Cultural layering of sacredarchaeological sites allows to study forms ofinheritance and to identify stages in the developmentof navigation technologies.
Development of instrumental navigation isconnected with development of the abstract thinking
and intelligence. The golden age of technology offore sight can be correlated to appearance of the
Homo sapiens and the Stone Age, and broad
application of a gnomon – to a metal era (about 7000
years ago).
When dating ancient objects it is necessary toconsider speeds of change of rocks and a relief ofregions of Earth for the last 2000000 years.

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Construction Material383

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ACCUMULATIVE DAMAGE OF FIBER SHEET REINFORCEMENT
CAUSED BY NEGATIVE THERMAL EXPANSION COEFFICIENTUNDER CYCLIC TEMPERATURE
1Hidenori TANAKA1
National Institute of Technology, Gunma College, Japan

ABSTRACTFiber sheet has been used in retrofitting of concrete structures after The Southern Hyogo Prefecture Earthquake
in 1995. It has advantages of light weight, high strength, high rigidity as well as easy to construct. However it
has negative thermal expansion coefficient that may cause shear stress between the interface fiber and epoxy
resin leading to de-lamination in interfacial area and deterioration of structural strength and durability of theconcrete structures. This paper presents experimental studies on interfacial damage of carbon fiber, aramid
fiber ,and super maintenance Hozen material using freeze-thaw process test at cyclic low temperature from 10℃ to 3℃, the typical average winter temperature of Hidaka City, Japan. It was found that at 6,000 cycles theplastic strains have been increased in aramid and super maintenance (Hozen material),however the accumulated
damages of those materials have been less than that of carbon fibers.
Keywords: Negative thermal expansion coefficient, Freeze-thaw test, Fiber sheet, Damage,Cyclic temperature

×100×400 (mm)). All specimens are shown in Table
1.,Carbon and aramid patterns in Figure 1.The
admixture of concrete and properties are shown in
Table 2 and Table 3, and then the properties of fiber
primer and Hozen material are shown in Table 4,
Table 5, Table 6 and Table 7 respectively.

1. INTRODUCTION
Concrete is a standard brittle material in civilengineering. Reinforced concrete structures withsteel bar resisting tension have been used forinfrastructure such as buildings, bridges and tunnels,
due to its durability and cost performance. In recent
years, special attention has been paid to fiber sheets
to resist against earthquake damage and prevent
separation failure [1]. The fiber sheet is of lightweight (1/5 of steel),high strength (10 times ofsteel),high rigidity and high durability as well aseasy to construct. However, the fibers have anegative thermal coefficient. So, there are someconcerned with mechanical interfacial properties
between concrete and fiber sheets, or epoxy resin
which glued fibers [2] - [4].
We have carried out some experiments with three
specimens by a freeze- thaw process test. According
to them, there were accumulation of plastic strain
and damage for the specimen with carbon fiber sheet
by 6,000 cyclic temperature which has been changed
from -10 to 3 centigrade. On the other hand, there
was little damage for the aramid fiber sheet and the
Hozen material that is one of coating methods bypolymer cement to reinforce the concrete structures.

Table 1 Specimens
SERIESTYPE

REMARK

CASE 1

BASE

CASE 2PRIMER

t=0.3 (mm)

CASE 3

CARBON FIBER

1 layer (t=0.22 (mm))

CASE 4

ARAMID FIBER

1 layer (t=0.42 (mm))

CASE 5HOZEN

t=10 (mm)t:thickness

carbonaramidFig.1 Specimens of fiber series (CASE 3,CASE4)
Base concrete

2. CYCLIC TEMPERATURE CHANGE TEST
Hozen material(t=10 mm)

2.1 Specimens
We applied the cyclic temperature change test for 5
specimens using a freeze-thaw process test machine.
The basic shape of the specimen is a prism(100Fig.2 Specimen of Hozen material384

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 Admixture of concrete
Gmax

Slump

(mm)

W/C

Air

Water
174

(%)
(%)
8
55
7
Unit value (kg/m3)
CementSand
Gravel
317
761
960

s/a

(cm)

15

maximum and minimum values (from -10 to 3
centigrade) in Hidaka City where the northernmost
concrete pier with carbon fiber was constructed.

(%)
44.5
S.P
3.17

Table 3 Properties of Concrete
Properties
value
Compressive strength 35 (N/mm2)
Poisson ratio
0.2
Young Modulus
30.2(kN/mm2)
Fig.3 Set on strain gauges for CASE2
Table 4 Properties of Carbon fiber
Properties
value
Unit weight308 (g/m3)
Tensile strength
4,657(N/mm2)
Young Modulus
265(kN/mm2)
Table 5 Properties of Aramid fiber
Properties
value
Unit weight
660 (g/m3)
Tensile strength 3,116(N/mm2)
Young Modulus
131(kN/mm2)
Table 6 Properties of primer
Properties
value
Compressive strength 43(N/mm2)
Adhesive strength
3.6(N/mm2)
Young modulus
2.45(kN/mm2)

Fig.4 Freeze-thaw test machine

Table 7 Properties of Hozen material
Properties
value
Bending strength12.3(N/mm2)
Compressive strength 55.3(N/mm2)
Adhesive strength
2.8(N/mm2)
Young modulus
18.7(kN/mm2)

Fig.5 Temperature change

These fibers and primer are standard materials to
repair and retrofit the concrete structures forexample ,tunnels, piers and box culverts.
2.2 Freeze-thaw test and principal strain
The strain gauges were respectively set on the centerand the corner of the specimens. The corner’s gauge
could be checked the principal value using byrosettes type which are arranged at 90 degree offsetsfrom each other.
The specimens were settled in the freeze-thaw test
machine (refer to Figure 4). And the temperaturechanges were determined considering the mean of

Fig.6 Curing condition
The temperature was changed regularly considering
the maximum and minimum range of the machine’s
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GEOMATE- Bangkok, Nov. 14-16, 2016

capacity (refer to Figure 5).
In this study, the freeze-thaw test was conducted in
air condition .There are generally two ways to set
the specimens in air or in water (refer to Figure 6). If
we set them in water, we couldn’t specify
deteriorated factors caused by the different thermal
expansion coefficient of fibers or the frost damage
by water. Therefore, we carried out the experiments
simply in air condition in order to make sure of theaffection by the different thermal expansion
coefficient of fibers. In addition, it has been known
that the diffusional coefficient of concrete in wet has
strong nonlinearity around 0 centigrade [5]. But, it isvery important and difficult to measure and evaluate
the nonlinearity on the surface of concrete. And
most of the structures in real environmental
condition haven’t been exposed in wet at all. So, in
this experiments, we focused on the properties offibers, and then we conducted simple condition.

Fig.8 Temperature-strain (ARAMID FIBER)

3. RESULT OF MEASUREMENT

CyclesCycles

3.1 Relationship between temperature andprincipal strain for specimens

CyclesCyclesCyclesCycles

The relationship between temperature and principal
strain of fiber series and Hozen material are shownin Figure 7 , Figure 8 and Figure 9 by 6,000th cycles.
The patters of primer and base in Figure 10 and
Figure 11 by 2,000th cycles.
Judging from the results of primer and base , they
have perfectly glued each other and there were notany damages among them. andthenthemeasurement of 2 specimens were carried out by
2,000th cycles.

Temp.

CyclesPrincipal strain

CyclesCyclesFig.9 Temperature-strain (HOZEN)

CyclesCyclesCyclesCyclesCyclesFig.7 Temperature-strain (CARBON FIBER)
Fig.10 Temperature-strain (PRIMER)
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GEOMATE- Bangkok, Nov. 14-16, 2016

have been occurred among fibers and epoxy resin[6] - [8].
4.DURABILITY INDEXES
I proposed durability indexes to focus on thehysteresis curves of carbon , aramid and Hozen
material corresponding to cyclic numbers.
4.1 Strain energy
At first, I checked the strain energy that is definedby area of hysteresis curves approximated ellipse(refer to Figure 12).

Fig.11 Temperature-strain (BASE)
Table 8 Representative strain of all specimens

CASE 4
CASE 5TYPE TEMPERATURESTRAIN （×0.000001)
BASE-9.5～-8.7 (℃) 258(10), 247(1000), 263(2000)
PRIMER -9.5～-8.7 (℃) 291(10), 313(1000), 281(2000)
CARBON -9.5～-8.7 (℃) 691(10), 843(1000), 1181(2000), 1421(3000)
1551(4000), 1630(5000), 1705(6000)
ARAMID -9.5～-8.7 (℃) 265(10), 408(1000), 441(2000),458(3000)
398(4000), 403(5000), 495(6000)
HOZEN -9.5～-8.7 (℃) 521(10), 882(1000), 928(2000),911(3000)
1163(4000), 1210(5000), 1431(6000)

Temp.

SERIESCASE 1
CASE 2
CASE 3

Datum
Approximated ellipse

Bracket means cyclic numbers
Representative measured strain values are shown in
Table 8.
The principal strain of the carbon fibers has beenchanged due to cyclic numbers. Especially, theplastic strain and the gradient of the loop curves
have been increased under cyclic temperature. The
inverse of gradient has shown the thermal expansion
coefficient of the specimens. And then, thisphenomena meant that fibers have split out theepoxy resin. Therefore, we considered that thedamage also has been accumulated between epoxy
resin and carbon fibers.
On the other hand, the principal strain of the aramid
fiber sheet scarcely has been changed due to cyclic
numbers. They were without damage under cyclic
temperature. In addition ,The specimens of primer
and base concrete have exhibited little difference inany cyclic temperature change. This meant primer
has perfectly glued the base concrete. The
accumulated damage depended on the fibers.
The plastic strain of Hozen material has been
increased due to cyclic numbers. But, the gradient ofloop curves has hardly been changed.
And then, the principal strains of primer and basehave been almost same curves under cyclic
temperature. So, primer has glued base concrete
perfectly by 2,000th cycles. The accumulative
damagePrincipal strain

Fig.12 Strain energy by approximated ellipse□ Hozen 1○ Hozen 2△ Carbon fiber

Strain energy

× Aramid fiber

Cyclic numbers

Fig.13 Relation strain energy and cyclic numbers
The values of strain energy about Hozen and carbon
fiber have been decreased due to cyclic numbers.
This means that the absorbable ability of materials
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GEOMATE- Bangkok, Nov. 14-16, 2016

corresponding to deformation have been decreased.
From the results, aramid fiber has no influence orsmall influence by temperature changes.

The thermal expansion coefficients of carbon fiber
have been decreased due to cyclic numbers. In thisstudy, Temperature increment ∆T has been constant(-10 to 3 deg.).But The thermal strain increment∆ε has been decreased. This means that the thermal
expansion coefficients α have been decreased by
the accumulative damage D ( D is damage variable).
Therefor, The thermal expansion coefficients α

4.2 Apparent thermal expansion coefficient
I secondly checked apparent thermal expansion
coefficient that approximated linear gradient ofhysteresis curves (refer to Figure 14).
Generally speaking, there is linear relationshipbetween thermal strain increment and temperature
increment using thermal expansion coefficient.

((1)∆ε = α∆T
Where, ∆ε: is thermal strain increment, α isthermal expansion coefficient , ∆T is temperature
increment.

5. CONCLUSION

Temp.

In this study, we introduced the experiments anddurability indexes under cyclic temperature changes.
The following conclusions have been obtained:(1) The plastic strain and damage of the carbon
fiber have been increased due to cyclic temperaturechanges by 6,000th . This damage may be occurredbecause of carbon fibers slipping out of epoxy resin.
(2) The strains of the aramid fiber have onlychanged from 100th to 1000th .We considered there
were not fatal damage under cyclic temperature.
(3) The strains of primer and base concrete havenot been changed by 2000th .This means the primer
perfectly glued the concrete as a result that the bothprincipal strain have been distributed nearly equal
under the cyclic changes.
(4) The strain energy has been decreased under thecyclic temperature change excepting the case ofaramid fibers. The damages among the laminatedmaterials have not been evaluated using these values.
(5) The thermal expansion coefficients of carbon
fibers have been decreased due to cyclic numbers.
These values have been closely correlated with theaccumulative damages between fibers and epoxy.
(6) We would like to carry out the analyses andexperiments about the carbon , the aramid fibers andHozen material block to evaluate final fatigue mode
under this cyclic temperature change .
(7) We have to consider more complicated natural
conditions (high humidity, high temperature change,
nonlinearity of diffusional coefficient) in next stage.

Datum
Approximated line

Principal strain

Thermal expansion coefficient ［×10-6／℃］

Fig.14 Thermal expansion coefficient byapproximated line

0.05

□

Hozen 1○

Hozen 2△

Carbon fiber×

Aramid fiber)

have been changed α 1 − D [9] , [10].
From the results of carbon fiber, The damages havebeen accumulated between fibers and epoxy resinthat glued fibers and primer.

0.04

0.03

6. ACKNOWLEDGMENT
0.02

The authors appreciate Nippon Steel Composite Co.,
Ltd. for providing the carbon and aramid fiber. andAssociation of Japan Construction Maintenance forproviding Hozen material and specimens. In addition,
This research was supported by J.S.P.S KAKENHI(No.24560572).

0.01
1000

2000

3000

4000

5000

6000

Cyclic numbers

Cyclic numbers

Fig.15 Relation thermal expansion coefficient andcyclic numbers
388

GEOMATE- Bangkok, Nov. 14-16, 2016
7. REFERENCE[1] Concrete
library101.”Repaire
andReinforcement guideline of concrete structureswith continuous fiber sheet,”,2000,pp.8-19.
[2] Zhishen Wu,Kentaro Iwashita.”Temperature
Effect on Bonding and Debonding Behavior
between FRP Sheets and Concrete,” J.Soc.Mat,.
Sci.,Japan,Vol.54,N0.5,2005,pp.474-480.
[3] Hidenori TANAKA,Yutaka TOI.”Adhesive
Failure Analysis of Structural Elements
Reinforced with Carbon Fiber Sheets, ”
ofTransaction of the Japan society
MechanicalEngineering,
SeriesAol.72,No.724,2006,pp.198-204.
[4] Tanaka H,Toi Y.,Maeda,K.,Sakai,T.” Damage
and fracture analysis of brittle structuralelements
reinforced with carbon fiber
sheets, ”JOURNAL OF ENVIRONMENTAND
ENGINEERING,Vol.3,No.1,2008,pp.111-122.
[5] Uekohara Akihisa,Sugawara Takashi.”Bonding
behavior of continuous fiber sheet under freezethaw,”Proceedings of the Japan Concrete
Institute,Vol.28,No.1,2006,pp.1715-1720.
[6] Makinoshima, F.,Tanaka H.”Influence on
Adhesion Characteristic of Concrete with F.R.P
Under Cyclical Temperature Change,”the 41st

JSCE
Kanto
BranchTechnical
presentation,2014, V-13.
[7] Tajima,R.,Tanaka H.”Deformation of Concrete
block with Super-hozen ,Fiber sheets under
Cyclic Temperature Change, ”the 42nd JSCE
Kanto Branch Technical presentation,2015, V42.
[8] Tanaka H.,Makinoshima F.,”The adhesive
Characteristic of Concrete Block with Fiber
Sheets Under Cyclic Temperature,”International Conference on the Regeneration
and Conservation of Concrete Structures
RCCS,2015,I-17.
[9] Lemaitre,J.,Desmorat.R..” Engineering Damage
Mechanics,” Springer,2005,pp.10-18.
[10] Tanaka H,Toi Y.”Ultimate Strength and Fatigue
Life of Concrete Members Reinforced with
Carbon Fiber Sheet Based on Continuum
DamageMechanics,”3rdInternational
Conference on Sustainable Construction
Materials&
Technologies–
SCMT3,2013,pp.99-108.

389

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

UTILIZATION OF AGGREGATE QUARRY WASTE INCONSTRUCTION INDUSTRY
Mary Ann Q. Adajar1, Euclid de Guzman2, Ryan Ho2, Cesar Palma Jr. III2, and Dennis Sindico2
1
Associate Professor, De La Salle University, Manila, Philippines
2
Civil Engineer, De La Salle University, Manila, Philippines

ABSTRACTThe use of solid wastes as a new ingredient in construction materials is one possible innovative effort to
reduce environmental degradation and to facilitate long term rational and sustainable use of the natural resources.
The disposal of the huge amount of wastes from aggregate quarry (WAQ) is one adverse environmental effectsof quarrying activities. Finding useful application of this solid waste, specifically as a substitute for fineaggregates in concrete mix, alleviates disposal problems and helps the construction industry to come up withconcrete products at lesser cost. This study investigates the structural performance of concrete with WAQ as
substitute for fine aggregates in a concrete mix in accordance with ASTM standards. Concrete with WAQ as fineaggregates achieved almost 78% of its target compressive strength at 28th day curing period. The reduced
compressive strength is due to the finer and less-angular particles of WAQ in comparison to sand. Empiricalmodel was formulated that can be used to predict the compressive strength of concrete with WAQ as substitutefor sand. Using the formulated model, the optimum compressive strength can be achieved at 85% substitution
for sand. The flexural strength of concrete with WAQ was in the range of 69% to 72% of the flexural strengthof concrete without WAQ. The presence of WAQ in the concrete mixture has no significant effect in its unit
weight, however, it affects the workability of the mix due to its cohesive property thus requires higher watercement ratio. Test results proved that concrete with WAQ as substitute for fine aggregates has strength
properties adequate for structural application.
Keywords: Aggregate Quarry Waste, Compressive Strength, Flexural Strength, Unit Weight

wastes as an alternative construction material.
Recycled aggregates produced from pulsed power
technology can be used as concrete component withfly ash as substitute for cement [1]. Waste carpet
fiber can suitably be used as fiber reinforcement in
concrete [2]. Tungsten mine waste with groundgranulated blast-furnace slag is found to be feasible
in the development of mortar with acceptable
compressive strength [3]. Paper mill sludge, anindustrial waste generated by paper mill factories isviable as partial replacement of fine aggregates inmanufacturing fresh concrete intended to be used for
low cost housing projects [4]. Ceramic waste andquarry dust aggregates have been found as a possible
replacement for conventional crushed stone coarse
and fine aggregates [5].
This study discusses the structural performanceof concrete with waste from aggregate quarry(WAQ) as substitute for fine aggregates in concrete
production.
Specifically, it determines thecompressive and flexural strength of concrete mix
when fine aggregates are partially and fully replacedwith WAQ in accordance with ACI 211. The study
describes the most suitable concrete mix proportion
with WAQ as fine aggregates that will yield thehighest strength of concrete in compression andflexure. It also investigates the effect of using WAQ

INTRODUCTION
Concrete is one of the most used building
materials. The boom in construction industry leads
to the exploitation of natural resources like riversand and gravel since these are the main source ofraw materials for concrete production. The onset ofshortage and increasing cost of construction
materials has made it important to consider
alternative materials that can be used in constructionindustry. Recycling waste to be potential materials
used in construction has become a trend worldwide
as this facilitates long term rational and sustainable
use of the natural resources and reduces construction
cost.
Quarrying activities generate huge amount ofsolid wastes which are most often dump in open
fields. The disposal of aggregate quarry waste is
one adverse environmental effect of quarrying
activities. Improper management of this wastematerial can bring serious detrimental effects to the
population and the environment. The use of solid
wastes as new ingredient in construction materials is
one possible innovative effort to reduce
environmental degradation and alleviate disposal
problems.
Several studies discussed the utilization of solid390

GEOMATE- Bangkok, Nov. 14-16, 2016

in the workability of concrete mix and in the unit
weight of hardened concrete.

for comparison of the obtained test results. The
target compressive strength of the concrete mix was
28 MPa. Compressive and flexural strengths were
measured after 7th , 14th, 21st, and 28th days curing
period designated as 7D, 14D, 21D, and 28D curingdays, respectively. Prior to testing, the specimens’
weight and volume were measured to determine theunit weight of hardened concrete.

WASTE FROM AGGREGATE QUARRY
Waste from aggregate quarry is a by-productgenerated from quarrying activities during the
production of aggregates. The solid wastes werecollected from a quarrying site at Sapang I, Ternate,
Cavite. These are waste products from crushing andwashing of mountain rocks to produce coarse andfine aggregates. The residues, which are considered
as solid wastes, were produced during the washingof crushed rocks in the siltation pond through thenatural process of sedimentation. Moist samples
from this source have grayish black color with softconsistency resembling that of fine sand when dry.
Previous studies discussed the geotechnicalcharacteristics of WAQ and was proven to besuitable embankment material [6,7].
Prior toexperimentation, the solid wastes were dried andsieved to ensure that the grain size conforms to the
ASTM requirements for fine aggregates.

Table 1 Concrete mix proportion
Mix Proportions
0B
25B
50B
75B
Sand
Vol. (m3)kg.
WAQ
Vol. (m3)kg.
Gravel (kg)Water(liter)Cement(bag)

EXPERIMENTAL PROGRAM
A series of laboratory tests in accordance to
ASTM standard procedures were carried out todetermine the structural performance of concretewith WAQ as substitute for fine aggregates. Grainsize analysis, Scanning Electron Microscopy (SEM)
test, and Energy Dispersive X-Ray Spectroscopy(EDS) were performed to determine the grain-size
distribution, microfabric structure and chemicalcomposition of WAQ, respectively. Its physical andchemical properties were analyzed and comparedwith that of conventional fine aggregates (sand).
To determine the compressive strength,
cylindrical specimens with dimensions of 150mm
diameter by 300mm high were subjected tocompression test in accordance to ASTM C39. To
determine the flexural strength, beam specimens
with dimensions of 150mm x 150mm x 500mm
were subjected to flexural strength test under a thirdpoint loading condition following the proceduredescribed in ASTM C78. A total of 92 cylindrical
specimens and 76 beam specimens were tested.
In this study, WAQ was used as partial and full
substitute for sand as fine aggregates in a concretemix. A typical concrete mix proportion is shown in
Table 1. The percentage substitution is done byvolume using a specific gravity (Gs) of 2.7 for sandand 2.57 for WAQ [6]. The slump of each mixture
was maintained at 2 to 3 inches. The 100B mixture
requires greater amount of water to achieve therequired slump and to make the mixture workable.
A conventional concrete without WAQ, designatedas 0B was used as control specimen and base data

100B

0.028
75.60

0.021
56.70

0.014
37.80

0.007
18.90

0.0
0.0

0.0
0.0

0.007
17.99

0.014
35.98

0.021
53.97

0.028
71.96

109.8

109.8

109.8

109.8

15

15

15

15

109.8
15 to
18

1

1

1

1

1

TEST RESULTS
Grain-Size Analysis
The distribution of grain sizes of WAQ andsand was determined using sieve analysis bymechanical method in accordance with ASTM D422.
Fig. 1 illustrates the grain size distribution curve.
The grain-size distribution curve shows that WAQ
has more percentage of finer particles as comparedto sand.

Sand
WAQ

Fig. 1 Grain-size distribution curve of WAQ andsand
The chemical composition of WAQ was
obtained from EDS test.
Table 2 presents the391

GEOMATE- Bangkok, Nov. 14-16, 2016

chemical composition of WAQ in comparison tosand. The data showed that WAQ and sandconsisted of the same chemical elements but withdifferent proportions. WAQ has 5.52% more silicon,
5.99% more Iron and 1.11% more calcium ascompared to sand. The bonding elements that mix
the aggregates together are composed of silicon,
calcium, and iron. Silicon is a significant element incement amounting to 17% to 25% of its chemicalcomposition [8]. Fresh concrete containing WAQ is
observed to be more cohesive and less prone tosegregation as compared to concrete mix without
WAQ. However, water requirement increases forthe concrete mix with more WAQ to improve its
workability. These observed effects are due to thepresence of more silicon content, similar to theeffect of mineral admixture like silica fume. The
presence of silica fume in concrete increases thedegree of flocculation, substantially reduces
bleeding by physically blocking the pores in the
fresh concrete and results to improved strength
properties [9].

At higher magnification (Fig. 2b), the micro
fabric of WAQ appears to be assemblages ofclustered platy particles while sand’s grain consistedof flaky particles arranged in some random direction
with large inter-assemblage pore spaces depicting
loose packing.

Table 2 Chemical composition of WAQ and sandElement
WAQ Sand*%%Magnesium (Mg)
0.59
2.06
Aluminum (Al)
8.15
6.33
Silicon (Si)24.43 18.91
Potassium (K)
1.00
0.43
Calcium (Ca)
2.77
1.66
Iron (Fe)6.92
0.93
Sodium (Na)
1.10
3.68
Total Carbon (C) and Oxygen (O) 55.41 34.00* from Gallardo & Adajar (2006) [4]A tabulated summary of the compressive
strength of concrete specimens with different
variations of WAQ can be seen in Table 3.

WAQ

Sand

Fig. 2b Micrographs at 5000X magnification
Compressive Strength

Table 3 Compressive strength of concrete withvarying percentage of WAQ
Average Compressive Strength (MPa)Curing
0B
25B
50B
75B
100B
Day
7D
21.07 11.68 10.02 17.18 12.62
14D
24.23 14.64 13.13 19.12 14.94
21D
26.28 15.88 14.70 21.51 18.95
28D
28.44 17.28 15.93 22.19 20.61

Microfabric Structures
The control specimen (0B) reached the target
compressive strength of 28 MPa at the 28th day
curing period. Concrete with WAQ (25B, 50B, 75B,
and 100B) have compressive strength lower than 0B.
This was observed in all curing periods (7D, 14D,
21D, and 28D). Of the mix with WAQ, the 75B mix
achieved the highest compressive strength andattained an almost 78% of the target strength while
the 50B mix has the least compressive strength
which is almost 56% of the target strength at 28th
day curing period.
Figure 3 shows the compressive strength against
the curing period of the specimens. It was observedthat the increase in compressive strength withrespect to curing period of concrete with WAQ
follows the same trend as the increase in strength ofthe control specimen. The compressive strength
increase ranges from 7% to 17% per week. The 75B
mixture can be considered to have early compressive

The micro-fabric structure of WAQ and sandwas obtained using the Scanning Electron
Microscope (SEM). As seen in Fig. 2a, WAQ has
smaller grain sizes than sand. Its structure consistedof a combination of rounded and sub-angular grains
with more silt-size grains while sand is comprised ofangular grains of uniform-size particles.

WAQ

Sand

Fig. 2a Micrographs at 50X magnification
392

GEOMATE- Bangkok, Nov. 14-16, 2016

strength development since the achieved strength at
7th day is almost 61% of the target strength.

increases the bonding resistance of the particles.
From test results, the positive effect of cohesivenesswas observed when the quantity of WAQ was morethan the quantity of sand. However, at 100B mix,
the cohesive property of WAQ produced negative
effect in terms of the workability of the concrete mix.
More water has to be added in the mix to attain thedesired slump and to make the mix more workable,
thus resulted to reduction in compressive strength.
Formulation of Empirical Model to Predict theCompressive Strength
The compressive strength for each percentage
WAQ substitution was presented in normalized formby dividing the experimental data with thecompressive strength of the control specimen. The
normalization of data was done to obtain a modelthat will predict the compressive strength of concrete
mix with WAQ substitution at any target
compressive strength and at any curing period.
Figure 5 shows the normalized graph of thecompressive strength against the percentage WAQ
substitution. The data points were plotted in ascattered matter and yielded a trend line equation in
the form:

Fig. 3 Compressive strength vs. curing period
Figure 4 shows the relationship between thecompressive strength and the WAQ substitution in a
concrete mix.

fc = (− 7.505 x 4 + 11.001x 3 − 2.176 x 2 − 1.656 x + 1) fco(1)
where:
fc = the compressive strength of concrete with x%
WAQ
fco = the compressive strength of concrete
without WAQ (control specimen)
x = WAQ substitution (in decimal form, i.e. 25%
is 0.25)

Fig. 4 Compressive strength vs. WAQ Substitution
The addition of WAQ resulted to a decrease incompressive strength. Comparing the compressive
strengths with that of 0B mix, the maximum
decrease in strength was observed in 50B mix while
the minimum decrease was noted in 75B mix. The
decrease in compressive strength can be attributed to
the finer particles of WAQ in comparison to sand as
seen from the grain-size distribution curve (Fig. 1).
Moreover, the shape of WAQ particle is rotund andless angular than sand as seen from SEM results (Fig.
2a and Fig. 2b). Finer and less angular particles arefactors that can contribute to reduction in particles’
frictional resistance, thus can lead to reduction incompressive strength [10].
However, when thepercentage of WAQ was more than 50%, it started togain a slight increase in compressive strength. When
WAQ was more predominant in quantity than sand,as in the case of 75B, it contributed positive results
which can be attributed to its cohesive property, thus

Fig. 5 Normalized Compressive Strength vs. WAQ
Substitution
The equation has a coefficient of determination,
R2 = 0.9631. The coefficient of determination, R2
393

GEOMATE- Bangkok, Nov. 14-16, 2016closed to 1.0 indicates that the model fits the datawell. From the equation obtained, the optimum
WAQ percent substitution for sand is at 85% inorder to produce the highest compressive strength ofthe concrete mix with WAQ.

The Modulus of Rupture of concrete represents
its flexural strength. It is the capacity of concrete toresist failure in bending. The average flexuralstrength of the specimens is presented in Table 4. It
was observed that the flexural strength of concretewith WAQ was in the range of 69% to 72% of theflexural strength of control specimen (0B). The
reduced flexural strength can be attributed to thesmooth surface texture of WAQ. Smooth texture of
aggregate results to weaker bond between theconcrete paste and the aggregates, thus leads tolower strength. Furthermore, the flexural strength of
concrete with WAQ was found to be 12% to 18% of
its compressive strength. This is in agreement withthe findings of National Ready Mixed Concrete
Association [12] that the flexural strength ofconventional concrete is about 10% to 20% of thecompressive strength.

Statistical Analysis on Compressive Strength
To verify the predictive capability of the
proposed model, the compressive strength (fc)calculated using the proposed model was comparedwith the measured values from experimentation.
Statistical analysis using T-test for paired samples
was performed to determine if there is a significant
difference between the measured values and thepredicted values. The paired T-test results for 16
number of observations with a level of significanceequals to 0.05 (α = 0.05) showed a t stat value (0.466)
less than t critical (1.753) and the p-value (0.324)
greater than 0.05. It can be stated, at 95%
confidence level, that there is no significant
difference between the measured compressive
strengths and the predicted compressive strengths
using the formulated model.
The strength ofassociation between measured fc and the predicted fc
was verified using the Pearson’s correlation
coefficient and the data are presented in a scatterplot as shown in Figure 6. The scatter of data points
is nearer to a straight line which means that there is alinear positive correlation between the measured
and predicted fc. The analysis yielded a Pearson
correlation value of 0.97 indicating a very strong
association between the two variables (measured andpredicted fc) [11]. It can be concluded that the
proposed model (Eq. 1) can be used to predict thecompressive strength of concrete with WAQ as
substitute for fine aggregates as a function of thecompressive strength of control specimen at any
curing period.

Table 4 Flexural Strength of Concrete with Varying
Percentage of Banlik
WAQ
FLEXURAL STRENGTH
SUBSTITUTION(MPa) AT 28TH DAY
0B
4.25
25B
2.98
50B
3.08
75B
2.94
100B
3.26
Unit Weight of Hardened Concrete with WAQ
The unit weight of hardened concrete is criticalto the performance of the structure. Concrete withlower density indicates the presence of more voids.
The grading and size distribution has a significant
contribution to the unit weight of concrete. It is
expected that when the particles are of uniform size,
the void spaces between particles are greater. But
when a varying range of sizes is used, the void
spaces are filled resulting to a larger unit weight.
The unit weight of hardened concrete was obtained
to determine the effect of WAQ substitution and theresults are presented in Table 5. The unit weight ofhardened concrete with WAQ was slightly lower
than the unit weight of 0B mix. The decrease in unit
weight was in the range of 1% to 2% which can beconsidered as insignificant. This was attested by the
result of statistical analysis. Statistical test of Oneway ANOVA yielded a P- value greater than 0.05.
This indicates that the presence of WAQ in theconcrete mixture did not produce significant
reduction nor increase in its unit weight.

Fig. 6 Correlation of measured fc with predicted fc

Table 5
KN/m3

Flexural Strength394

Unit Weight of Concrete Cylinders in

GEOMATE- Bangkok, Nov. 14-16, 2016

Curing
Day
7D

0B

25B

50B

75B

23.40

23.13

23.34

23.23

23.16

14D

23.37

22.9

23.09

22.99

22.93

21D

23.28

23.10

23.29

23.18

23.13

28D

23.92

23.34

23.54

23.44

23.37

REFERENCES

100B[1] Arifi E, Zacoeb A, and Shigeishi M, “Effect of
fly ash on the strength of concrete made fromrecycled aggregate by pulsed power”, Int. J. of
GEOMATE, Vol. 7, No. 1 (Sl. No. 13), Japan,
Sept. 2014, pp. 1009-1016.
[2] Abdul Awal ASM, Mohammadhosseini H, andHossain MZ, “Strength, modulus of elasticity
and shrinkage behavior of concrete containing
waste carpet fiber”, Int. J. of GEOMATE, , Vol.
9, No. 1 (Sl. No. 17), Sept. 2015, pp. 1441-1446.
[3] Yun Wang Choi, et.al, “Utilization of tailings
from tungsten mine waste as a substitution
material for cement”, Construction and Building
Materials, Vol. 23, 2009, pp. 2481–2486.
[4] Gallardo R. andAdajar M, “Structural
performance of concrete with paper sludge as
fine aggregates partial replacement enhancedwith admixtures.” JSPS-DOST Int. Symposium
on Environmental Engineering and 5th Regional
Symposium on Infrastructure Devt., 2006.
[5] Abdullah, M. et.al. “Concrete with ceramic
wastes and quarry dust aggregates.” 5th Annual
Conference Management in Construction
Researchers Association, 2006, pp. 383 – 388.
[6] Adajar, MQ and Zarco, MH, “An empiricalmodel for predicting hydraulic conductivity ofmine tailings”, Int. J. of GEOMATE, Vol. 7, No.
2 (Sl. No. 14), Japan, Dec., 2014, pp. 1054-1061.
[7] Adajar, MQ and Zarco, MH, “Predicting thestress-strain behavior of mine tailing using
modified hyperbolic model”,
Int.
J. of
GEOMATE, Vol. 10, Issue 21, Japan, May
2016, pp. 1834-1841.
[8] “Compositionofportlandcement”,
http://theconstructor.org/building/compositionof-portland-cement/5725/, Accessed : May 4,
2016.
[9] Rasol MA, “Effect of silica fume on concrete
properties and advantages for Kurdistan Region,
Iraq, Int. J. of Scientific & Engineering
Research, Volume 6, Issue 1, January 2015, pp.
170 – 173.
[10] Shetty M S, Concrete Technology Theory andPractice, Ram Nagar, New Delhi: S. Chand &Company Ltd., 2005.
[11] Evans JD, Straightforward Statistics for the
Behavioral Sciences, Pacific Grove, Calif.:
Brooks/Cole Publishing, 1996.
[12] National Ready Mixed Concrete Association,“Concrete in Practice CIP 16 – Flexural
Strength
Concrete.”,from
http://www.nrmca.org . Accessed: Jan. 29,
2015.

CONCLUSION
The structural performance of concrete mixed
with waste from aggregate quarry (WAQ) as a
substitute for sand was investigated. The following
conclusions were drawn from the experimental
results:
Waste from aggregate quarry contains moresilicon, iron, and calcium as compared to sand.
These are elements that caused concrete mixture
with WAQ to have an increased cohesive property.
SEM results showed that WAQ has finer, lessangular particles than sand. The finer particles of
WAQ and presence of more Silicon, Iron, andCalcium elements produced positive effect in termsof cohesive property of the concrete.
Concrete with WAQ as fine aggregatesachieved an almost 56% to 78% of the compressive
strength of concrete without WAQ. Empiricalmodel was formulated that can be used to predict thecompressive strength of concrete with WAQ as
substitute for sand. Using the formulated model, thehighest compressive strength can be achieved at
85% WAQ substitution for sand. The flexuralstrength of concrete with WAQ was in the range of
69% to 72% of the flexural strength of concrete
without WAQ and was found to be 12% to 18% of
its compressive strength. The reduced compressive
and flexural strengths can be attributed to the finer,
less-angular particles and smooth surface texture of
WAQ in comparison to sandThe presence of WAQ in the concrete mix has
no significant effect on the unit weight of hardened
concrete, however, it affects the workability of themix due to its cohesive property thus requires higher
water-cement ratio.
Test results proved that waste from aggregate
quarry can be used as substitute for fine aggregatesin a concrete mix and produced strength properties
adequate for structural application. This study offers
an effective waste management solution for WAQ
and introduces an alternative material in concrete
production thus contributing to a sustainable
development.

395

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

SPENT COFFEE GROUND-FLY ASH GEOPOLYMER: STRENGTHANALYSIS AS A RECYCLED ROAD SUBGRADE MATERIAL
TA. Kua1, A. Arulrajah2, S. Horpibulsuk3 and YJ. Du4
Department of Civil and Construction Engineering, Swinburne University of Technology, Melbourne,
Australia; 3 Suranaree University of Technology, Nakhon Ratchasima, Thailand; 4 Institute of GeotechnicalEngineering, Southeast University, Nanjing, China.
1,2

ABSTRACT
Coffee is a staple drink in many countries and cultures. The residue generated from coffee brewing, spentcoffee ground (CG), is an organic waste which sees little use apart from occasionally being recycled
domestically as an agricultural fertilizer. It is granular and physically similar to sand, hence the potential to bestabilized into a fill material for construction purposes. By combining CG with alkaline-activated Fly Ash (FA),a waste product generated as a result of coal combustion, a geopolymer was formed. An alkaline activator
composed of a sodium silicate+ sodium hydroxide (Na 2 SiO 3 + NaOH) mix was used to induce
geopolymerization. Strength assessment of this geopolymer was done via Unconfined Compressive Strength(UCS) tests. Factors found to affect the strength development of the CG-FA geopolymer are: (1) the ratio of
sodium hydroxide and sodium silicate in the activator liquid; (2) the curing time; (3) the replacement ratio of FA
in the CG; (4) the alkalinity of the activator liquid used; and (5) the curing temperature. Four-days soaked
California Bearing Ratio (CBR) tests conducted on selected mixes meet the requirement for a load-bearing
subgrade fill material. This study shows that highly organic CG can be reprocessed to meet strength
requirements as a recycled road subgrade material. The resultant product, which constitutes largely of solid
waste materials, can potentially be developed into a green industrial fill material and reduce landfill spacedemand.
Keywords: Spent-coffee-ground; Fly-ash; Geopolymer; Subgrade; Reycling.

due to its high organic content of 86-89% and low
load bearing strength.
Stabilizing CG for construction uses is a notionthat would ideally divert a steadily generated organic
waste into the construction industry, thus reducing
both landfill space consumption and raw material
demand. Chemical stabilization on soft soils
conventionally utilizes Portland cement and lime [9].
The treatment results in augmented soil strength andstiffness. However, these additives are purposefully
manufactured and are known to leave large carbon
footprints [10], hence are contradictory to the idea ofsustainability.
An alternative stabilization method bygeopolymerization may be used to treat CG to forma green construction material. Geopolymerization isa chemical process which generates cementitious
compounds but leaves minimal carbon footprints
compared to the production and utilization ofPortland cement [11]. Fundementally, the formationof cementitious compounds in geopolymers relies on
the co-polymerization of alumina and silica in thebase material, or precursor, by highly alkaline
solutions. Fly ash (FA) is an aluminosilicate-richmaterial and a well-established precursor frequently
used in geopolymerization [12]. It is a by-product ofcoal combustion and is easily obtainable due toglobal reliance on coal-combustion to generate

INTRODUCTION
Sustainability is a key issue in current researchesin construction technology. As a result, reprocessing
waste materials into recycled construction materials
have become a popular field of research. Wastes
including construction and demolition materials [1],[2], crushed glass [3], water treatment sludge [4] andhighly organic waste-water biosolids [5] have beentreated and laboratory-tested for their suitability as
recycled
construction
materials.
Successful
implementation of recycled waste materials in theconstruction industry would diminish the need toharvest raw materials for construction and reduce
dependency on landfills for waste disposal.
Spent coffee ground (CG) is the insoluble byproduct resulting from the thermal water extraction
from ground coffee beans. The global food andbeverage industry generates approximately 7.4
million tons of CG each year [6], and this figure is
expected to steadily increase due the popularity of
the coffee drink [7]. CG is commonly used as adomestic agricultural fertilizer, but otherwise nolarge scale utilization of this organic waste has beenreported. An analysis was recently done on CG [8]to assess its geotechnical properties and it wasreported that CG physically resembles sandy soils,
but is unsuitable for use as a construction material
396

GEOMATE- Bangkok, Nov. 14-16, 2016

50:50 were used at varying liquid-to-activator(L/FA) contents. The concentration of NaOH in
Stage 1was fixed at 8 mol. Subsequently, optimum
liquid contents (OLC) under modified compactionenergy was determined according to ASTM D155712. After OLC values were obtained, UCS samples
were prepared by varying L to incorporate samples
from both the dry and wet sides of the OLCs. Twosets of triplicate specimens were produced and cured
at 7 and 28 days separately. The curing temperatureused at this stage is 50°C, the maximum working
temperature for CG as described previously.
Stage 2 fixed the Na 2 SiO 3 :NaOH ratio at 90:10
and L/FA ratio at 1.8, as 90:10 specimens produced
the minimum UCS values among three
Na 2 SiO 3 :NaOH ratios in Stage 1. Also, from theresults of Stage 1, these 90:10 specimens achieved
maximum UCS at L/FA= 1.8 hence the L/FA ratio
used. Remaining controlled variables were tested asfollow: 1) The FA content was varied from 10%30% relative to the dry weight of CG, with NaOH
concentration fixed at 8 mol and curing temperature
fixed at 50°C; 2) NaOH concentration was varied at
5, 8, 10, 12, and 15 mol, with FA content fixed at
30%; and lastly 3) Curing temperature was varied
from 21°C (room temperature), 40°C (summer
temperature), to 50°C (maximum workabletemperature), with FA content fixed at 30% and
NaOH concentration fixed at 8 mol. All specimens
in the 2nd stage were cured for a 7-day period. The
Na 2 SiO 3 :NaOH ratio of 90:10 was selected to assess
the maximum achievable strength development
when NaOH content is at a minimum.
UCS samples were prepared by 1-layer static
compression [4],[17] in a 50 mm diameter by 100
mm height split-mould to reach the target dry
density previously obtained from the modified
Proctor compaction method. The specimens werewrapped and sealed in transparent vinyl wraps toprevent moisture lost, and cured in temperatureregulated ovens. Figure 1 illustrates the apparatus
used to produce the UCS specimens and also thewrapped specimens before curing.

energy [13]. Sodium hydroxide (NaOH) and sodium
silicate (Na 2 SiO 3 ) are the typical alkaline activators
used in geopolymerization [12], [14].
Melbourne, Australia, is a city known for its
deep-rooted coffee culture. On the other hand, a
study in 2005 shows that 75% of Australia’s energy
comes from coal combustion hence 12 million
tonnes of FA can be collected annually [15]. This
study aims to combine two locally available wastematerials, namely CG and FA, to form a green
construction material via geopolymerization.
Different ratios of NaOH and Na 2 SiO 3 , NaOH
molarities, curing durations, and curing temperature
were used to study the factors influencing strength
development in this hybrid material. Variations inmaterial strength due to these factors were observed
via Unconfined Compressive Strength (UCS) tests,
whereas the suitability of the material to be used as aroad subgrade fill material was determined by
California Bearing Ratio (CBR) tests.
MATERIALS AND METHODSCG was collected daily from a café in Melbourne,
Australia and oven dried at a temperature of 50°C toprevent the charring of organics and subsequent lossof mass at higher temperatures [8]. Due to therelatively low drying temperature and high moisture
content in CG, drying typically requires 4 to 5 days.
Moisture content was measured according to ASTM
D2216-98. Coagulated clumps of CG can becommonly found in the dried material, but can beeasily crumbled with abrasion by hand, indicating no
chemical bonding. Particle size analysis on dried
CG was done according to D442-63.
Commercially available Class F FA was used as
the precursor. FA categorized in Class F are low inlime content, hence does not react readily with waterand needs alkaline activation to show cementitious
properties. Combinations of NaOH and Na 2 SiO 3
were used as the alkaline activator. Commerciallyavailable reagent grade NaOH with 97% purity andD-grade Na 2 SiO 3 composed of 44.1% silicic acid,sodium salt (1.6<Molar Ratio≤2.6) and 55.9% water
were used. To prepare each geopolymer specimen,
dried CG was manually mixed with the required
percentage of FA in a bucket for 5 minutes to ensurematerial homogeneity. The required percentage ofliquid activator was then added to the dry material.
This mixture was further mixed for 5 minutes
manually. The mixture was then used immediately
for compaction to avoid excessive flocculation andhardening.
The UCS tests in this study, summarized inTable 1, consisted of 2 stages. In Stage 1, the ratio of
CG:FA was fixed at 70:30 because a 30% binder
ratio is known to successfully strengthen waste
treatment sludge, another organic material [16].
Three ratios of Na 2 SiO 3 :NaOH at 90:10, 70:30, andFig. 1 Photos of (a) 50 mm diameter by 100 mm
height split mould used for compaction; (b)apparatus setup during static compaction process; (c)397

GEOMATE- Bangkok, Nov. 14-16, 2016

CG geopolymer specimen after compaction, before
extrusion from split mould; and (d) extruded CG
geopolymer specimens sealed in vinyl.

suitability of an embankment fill or subgrade
materials [18]. These specimens were soaked for 4days immediately after compaction, to simulate
severe flooding of pavements in the open world
environment. 3 specimens were prepared for eachmaterial combination, with the final assigned CBRtaken as the average of the 2 lowest CBR results outof the 3 specimens.

CBR tests were done on selected mixes after
analyzing the UCS results. The CBR specimens
were prepared and tested in accordance with ASTM
D1883-07, as the CBR value is one of the key
criteria used by the local road authority to assess the
Table 1 UCS Testing Programme SummaryStage

Modified Variable

FA Replacement % by

Na 2 SiO 3 :NaOH

Dry Weight of CG

Ratio

(%)

(by weight)

Na 2 SiO 3 :NaOH
1

Ratio &

L/FA
(by weight)

NaOH

CuringMolarity

Temperature

Curing Time

(mol)

(°C)

(days)

1.5, 1.8, 2.2,
30

90:10

2.6

ActivatorLiquid

8

50

7,28

8

50

7,28

1.2, 1.5, 1.8,

Content

70:30

2.1
1.2, 1.5, 1.8,

50:50

2.1

8

50

7,28

10,15,20,25,30

90:10

1.8

8

50

7

30

90:10

1.8

15

50

7

30

90:10

1.8

8

21, 40, 50

7

FA Replacement
2

Ratio

NaOH
Concentration

5, 8, 10, 12,CuringTemperature

The compaction curves of 70:30 CG:FA mixes
are shown in Fig. 3 and 4. Figure 3 compares theprescribed amount of aqueous Na 2 SiO 3 :NaOH to
the prescribed dry weight of the CG:FA mix. On theother hand, Fig. 4 compares the amount of
evaporated water to the dry bulk density ofremaining solid, including the solid Na 2 SiO 3 :NaOH
trapped in the mix after evaporation. in the specimen
after drying.

RESULTS AND DISCUSSIONS
Figure 2 shows the particle size distribution
curve of dried CG. With particles in the range of
0.075 to 2.36 mm in size, CG resembles poorlygraded sand according to the classification method
in ASTM D2487-10, but due to the presence of highorganic content it should be identified as an organic
clay (OH).

Fig. 3 Dry density vs. liquid alkali activator content
for various Na 2 SiO 3 :NaOH ratios.

Fig. 2 Particle size distribution of CG.

398

GEOMATE- Bangkok, Nov. 14-16, 2016

hypothesized weaker geopolymer due to the lowamount of NaOH present in this mix.
The UCS values of geopolymer specimens inStage 1, namely with 30% FA cured at 50°C after 7
and 28 days, are illustrated in Fig. 5 and 6,
respectively. By analyzing Fig. 5, the maximum 7
days UCS is found at L/FA ratio of 1.8 for
Na 2 SiO 3 :NaOH ratios of 70:30 and 50:50 while it
was found at L/FA of 2.3 for the Na 2 SiO 3 :NaOH
ratio of 90:10. While 70:30 and 50:50 samples
achieved maximum UCS at OLC, 90:10 samples
require liquid contents exceeding its OLC. At L/FA
= 1.8 mixes with a 50:50 Na 2 SiO 3 :NaOH content
result in the highest strength at 956.3 kPa. However,maximum 7 days UCS strength was registered by
90:10 specimens at L/FA= 2.2, at 1160.5 kPa. On
the other hand, for an extended curing time of 28days as shown Fig. 6, the maximum UCS for 70:30
occurs at L/FA = 1.8, while 50:50 achieved
maximum UCS at L/FA = 1.5. In the case of 90:10,
a decrease in gradient starts to develop in the graphbeyond L/FA = 1.8, indicating a saturation point. In
all, UCS was found to increase when curing time
was extended from 7 days to 28 days.

Fig. 4 Dry density vs. moisture content for various
Na 2 SiO 3 :NaOH ratios.
From Fig. 3, it can be observed that the mixes are
relatively insensitive to moisture changes below
50% (with variations not exceeding 0.5 kN/m3),
compared to the sharp decrease in dry density as theliquid content is increased after the peak at 52%.
The 50:50 Na 2 SiO 3 :NaOH ratio specimens register
the highest maximum dry density (MDD) (8.6
kN/m3) relative to specimens compacted with theother ratios of Na 2 SiO 3 :NaOH (8.5kN/m3).
However, ASTM 1557-12 allows a measurement
error of approximately ±0.1 kN/m3 hence making
the difference barely significant. This MDD
achieved by CG:FA geopolymers is double that of
unstabilized CG, which has a reported MDD of 4.4
kN/m3 [8]. The MDD of these CG:FA geopolymers
are similar to that of compacted biosolids, but still
are relatively low compared to natural occurringsoils. Interestingly, the dry densities in Fig. 3 show
sign of increase as the amount of administered liquid
approaches 0%, resulting in a sinusoidal curve
particularly in the 90:10 Na 2 SiO 3 :NaOH mix. Thisoccurrence is typically observed in sands [19] where
MDD may occur at 0% moisture, but in this caseassigning an MDD of 0% to the mixes would bemeaningless because the liquid activator is needed
for geopolymerization to occur.
On the other hand Fig. 4 shows that all threemixes exhibit a similar dry bulk density of 10.5
kN/m3. As NaOH increases in the Na 2 SiO 3 :NaOH
ratio used, the amount of moisture corresponding to
the MDDs in the curves in Fig. 4 increases. This is
attributed to the high water content found in the
NaOH.
Strength evaluation of the geopolymers was doneby UCS tests. While the OLC was determined to be
52% from Fig. 3, L/FA ratios from both the wet anddry sides of the OLC at smaller intervals wereselected to pinpoint the maximum achievable UCS
relative to changes in material density. While L/FA=
1.2, 1.5, 1.3, 2.1 were used for the 70:30 and 50:50
Na 2 SiO 3 :NaOH specimens, a higher L/FA ratio wasnominated for 90:10 specimens to offset theFig. 5 7 days UCS versus liquid content for various
Na 2 SiO 3 :NaOH ratios.

Fig. 6 28 days UCS vs. liquid content for various
Na 2 SiO 3 :NaOH ratios.
As NaOH increases, the rate of precipitation andpoly-condensation of FA is accelerated [4], hence399

GEOMATE- Bangkok, Nov. 14-16, 2016the high strength generated by 50:50 specimens at
L/FA= 1.8 at 7 days curing. On the other hand, when
NaOH is excessively low, FA particlces cannot
dissolve efficiently [4], hence the need for higher
liquid contents in 90:10 mixes. Consequently, highlevels of silicate in a mix would also result in a highrate of geopolymerization [20] via the formation of adifferent silica-rich sodium aluminosilicate hydrate(NASH) gel hence the high 7-days UCS for 90:10
specimens at L/FA= 2.2. Comparatively, theDepartment of Highways, Thailand nominates aminimum subgrade UCS of 294.2 kPa [21], whereas
subgrade materials used in low-volume-traffic roads
in Malaysia are required to have a minimum UCS of
800 kPa [22]. 50:50 specimens at L/FA= 1.8 cured
for 7 days meet both benchmarks with the leastliquid required.
The 50:50 Na 2 SiO 3 : NaOH mixture at L/FA =1.8 was consequently selected to undergo 4-days
soaked CBR tests. Seperately, 90:10 and 70:30 CBR
specimens were produce at L/FA = 1.8 to obtaincomparable results. Figure 7 shows that 50:50
samples achieves the highest CBR strength at 12%.
Based on road authority specifications in the state ofVictoria, Australia, all specimens pass the minimum
acceptance requirement of 2% [18] to function asembankment structural fills or subgrade materials.

elevated temperature curing at 40°C greatly
improves UCS compared to samples cured at 21°C,
but increasing the curing temperature from 40°C to
50°C produces only a relatively small increase in
UCS.

Fig. 8 7 days UCS vs. FA replacement ratio, with
8mol NaOH, and 50°C curing.

Fig. 9 7 days UCS vs. NaOH molarity, with 30% FA
replacement, and 50°C curing.

Fig. 7 4 days soaked CBR values various
Na 2 SiO 2 :NaOH ratios.
Subsequently, 90:10 Na 2 SiO 3 :NaOH with
L/FA= 1.8 and 7 days curing was fixed and used to
study the effects of various FA contents, NaOH
concentrations, and curing temperatures on UCS
development. Figure 8 shows that at a curing
temperature of 50°C, UCS increases with FA
replacement ratio relative to CG, but after 25% thegradient decreases, implying a saturation point.
Figure 9 demonstrates that at a fixed FA content of
30% and curing temperature of 50°C, the UCS
increases as the molarity of NaOH used increases.
Optimally the molarity used should be in the rangeof 8- 12 mol where small increments in molarity
results in larger increase in UCS relative to otherparts of the graph. Lastly, Fig. 10 demonstrates that

Fig. 10 7 days UCS vs. curing temperature, with
8mol NaOH, and 30% FA replacement.
CONCLUSION
This research indicates that CG can be stabilized via
geopolymerization to produce a material withincreased compressive strength. Road authority
specifications in Victoria, Australia dictate that CBR
400

GEOMATE- Bangkok, Nov. 14-16, 2016

values for subgrade materials must exceed 2%. This
criterion is met by all the mixes studied. Specifically,
the mix of 70:30 CG:FA with 50:50 Na 2 SiO 3 :NaOH
and L/FA= 1.8 was found to be an efficient mix
when cured at 50°C for 7 days. tests. Factors found
to affect the strength development of the CG-FA
geopolymer are: (1) the ratio of sodium hydroxideand sodium silicate in the activator liquid; (2) the
curing time; (3) the replacement ratio of FA in theCG; (4) the alkalinity of the activator liquid used;and (5) the curing temperature. A recommendation
for future works would be to look into the long term
durability of this material to produce a green
construction material with acceptable design life.

[10] Horpibulsuk S, Phetchuay C, Chinkulkijniwat
A, Cholaphatsorn A. Strength development in silty
clay stabilized with calcium carbide residue and flyash. Soils and Foundations. 2013;53(4):477-86.
[11] Hicks J. Durable "Green" Concrete from
Activated Pozzolan Cement. Green Streets andHighways 20102010. p. 408-30.
[12] Sukmak P, Horpibulsuk S, Shen S-L. Strength
development in clay–fly ash geopolymer.
Construction and Building Materials. 2013;40:56674.
[13] Malhotra V. Role of FA in reducing greenhousegas emissions during the manufacturing of Portland
cement clinker. 2nd International Conference on
Advances in Concrete Technologies in the Middle
East Dubai2008.
[14]
PhetchuayC,HorpibulsukS,
Suksiripattanapong C, Chinkulkijniwat A, Arulrajah
A, Disfani MM. Calcium carbide residue: Alkaline
activator for clay–fly ash geopolymer. Construction
and Building Materials. 2014;69:285-94.
[15] Morrison AG, P. Nelson, P. Future fly ash
availabilityPotential
consequencesoftransformation of Australia’s energy generation
portfolio to 2050. World of Coal Ash (WOCA).
Lexington, Kentucky, USA.2005.
[16] Horpibulsuk S, Suksiripattanapong C,
Samingthong W, Rachan R, Arulrajah A. Durability
against Wetting–Drying Cycles of Water Treatment
Sludge–Fly Ash Geopolymer and Water Treatment
Sludge–Cement and Silty Clay–Cement Systems.
Journal of Materials in Civil Engineering.
2015:04015078.
[17] Sukmak P, Horpibulsuk S, Shen S-L,
Chindaprasirt P, Suksiripattanapong C. Factors
influencing strength development in clay–fly ash
geopolymer. Construction and Building Materials.
2013;47:1125-36.
[18] Vicroads. RC 500.22: Selection and Design ofPavements and Surfacings. Victoria, Australia2013.
[19] Rollings MP, Rollings RS. GeotechnicalMaterials in Construction: McGraw-Hill; 1996.
[20] Ravikumar D, Neithalath N. Effects of activator
characteristics on the reaction product formation inslag binders activated using alkali silicate powder
and NaOH. Cement and Concrete Composites.
2012;34(7):809-18.
[21] Teerawattanasuk C, Voottipruex P, Horpibulsuk
S. Mix design charts for lightweight cellular
cemented Bangkok clay. Applied Clay Science.
2015;104:318-23.
[22] Awg Shahminan DNIA, Rashid ASA, Ridzuan
Bunawan A, Yaacob H, Noor NM. Relationshipbetween strength and liquidity index of cement
stabilized laterite for subgrade application.
International Journal of Soil Science. 2014;9(1):1621.

REFERENCES[1] Arulrajah A, Disfani MM, Horpibulsuk S,
Suksiripattanapong C, Prongmanee N. Physicalproperties and shear strength responses of recycledconstruction and demolition materials in unboundpavement base/subbase applications. Construction
and Building Materials. 2014;58:245-57.
[2] Rahman MA, Imteaz M, Arulrajah A, Disfani
MM. Suitability of recycled construction anddemolition aggregates as alternative pipe backfillingmaterials. Journal of Cleaner Production.
2014;66:75-84.
[3] Disfani MM, Arulrajah A, Bo MW, Sivakugan N.
Environmental risks of using recycled crushed glassin road applications. Journal of Cleaner Production.
2012;20(1):170-9.
[4] Suksiripattanapong C, Horpibulsuk S,
Chanprasert P, Sukmak P, Arulrajah A. Compressive
strength development in fly ash geopolymer
masonry units manufactured from water treatment
sludge. Construction and Building Materials.
2015;82:20-30.
[5] Arulrajah A, Disfani MM, Suthagaran V, Bo
MW. Laboratory evaluation of the geotechnicalcharacteristics of wastewater biosolids in roadembankments. Journal of Materials in Civil
Engineering. 2013;25(11):1682-91.
[6] Kondamudi N, Mohapatra SK, Misra M. Spent
Coffee Grounds as a Versatile Source of Green
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2008;56(24):11757-60.
[7] Butt MS, Sultan MT. Coffee and its
consumption: benefits and risks. Critical reviews infood science and nutrition. 2011;51(4):363-73.
[8] Arulrajah A, Maghoolpilehrood F, Disfani MM,
Horpibulsuk S. Spent coffee grounds as a nonstructural embankment fill material: Engineering andenvironmental considerations. Journal of Cleaner
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401Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

OPTIMIZATION OF COMPRESSIVE STRENGTH OF CONCRETE
WITH PIG-HAIR FIBERS AS FIBER REINFORCEMENT AND GREENMUSSEL SHELLS AS PARTIAL CEMENT SUBSTITUTE
1,2

Engr. Jayvee L. Gagan1 and Dr. Bernardo A. Lejano2
Civil Engineering Department, Gokongwei College of Engineering
De La Salle University, Philippines

ABSTRACTThe feasibility of different waste materials as substitute to the main components of concrete is attracting
attention nowadays. In relation to that, this study focuses on determining the effects of combining two wastematerials namely, pig-hair fibers (PHF) as fiber reinforcement and crushed green mussel shells (GMS) as
partial cement substitute to the properties of concrete. Response Surface Methodology (RSM) was used tomodel the relationship between the response and the factors considered. Using central composite design (CCD)
to establish the design of experiment, the researchers was able to reduce the required number of experimental
runs to 20 from a total of 27 runs for 3 level full factorial experiment that is common for responses withnonlinear behavior. Optimization was conducted to determine the optimum amount of PHF and GMS in
concrete that could yield maximum compressive strength while keeping the workability at an acceptable level.
As for the results, an increase in compressive strength of concrete was recorded with the incorporation of PHF
and GMS to concrete. However, decrease in workability was experienced due to the amount of fiber
reinforcement present in the mix. Results of RSM suggested an optimum combination of 0.70% PHF content
and 7.81% GMS partial cement substitute at 0.47 w/c ratio to achieve 27.40 MPa and 2.78 MPa compressive
and tensile strength respectively with a minimum recommended slump of 25 mm for concrete beams andcolumns as per ACI. Based on these results, PHF-GMS concrete could be used in structures not requiring
compressive strength above 28 MPa and with the use of GMS as a partial cement substitute, it could reduce theoverall cement requirement for a project thus incurring savings and most importantly promotes the use ofenvironment friendly materials.

Keywords: Pig-hair Fibers, Green Mussel Shells, Response Surface Methodology, Concrete
cement replacement are examples of application ofgreen concrete technology.
Generally, materials with high lime or silica
content could be a potential partial substitute to
cement. In relation to this, most sea shells composedof calcium carbonate or limestone that made up
almost 95% of its composition. Due to the presenceof lime, shells exhibit cementitious characteristics
when subjected to refinement. A chemical analysis
was conducted by Etuk [3] to determine thechemical composition of powdered shells and basedon the results, shells contains large amount of CaO(calcium oxide) and SiO 2 (silica) that are the twomain chemical compounds in ordinary Portland
cement making it feasible as partial cement
substitute. Scallop shell (SS) powder was used as
cement additive for grouting, results showed that theunconfined compressive strength of sand piece
samples increased by 25% (120 kPa to 150 kPa)
upon incorporation of 10% SS powder [4]. Also, a
study was conducted on the utilization of Green
Mussel Shells (GMS) and results showed thatincorporating GMS into concrete as partial cement
substitute improves its compressive and tensile

INTRODUCTION
Concrete is considered as the second mostconsumed substance in the world next to water. As
of 2012, a total of 11.5 billion tons of concrete areproduced each year and is expected to increase toapproximately 18 billion tons a year by 2050 [1].
Based on a study by Rubenstein [2], approximately
5% of the total carbon dioxide emission all over theworld is constituted to concrete production. Asidefrom pollution, another problem on the increasing
demand for concrete production is the shortage on
locally available raw materials especially in remoteareas.
Due to the problems stated above, someresearchers [3], [4], [5] today are inclined to study
the feasibility of using other materials as substituteto the main components of concrete i.e. cement, sandand gravel. These previous studies were developed
to help minimize the negative effects of concrete
production on the environment and provide cheaper
alternative for the consumers. Addition of natural
fiber reinforcement to improve strength and partial402

GEOMATE-Bangkok, Nov. 14-16, 2016

strength by 48.28% and 68.06% respectively [5].
These studies illustrated the potential of limestone
from sea shells as substitute cementitious material
for concrete.
Fiber reinforcement in concrete is used toprovide additional durability and improve its overall
performance. Polymer based synthetic fibers are theones that are commonly used in the constructionindustry as reinforcement. They can be classified as
microsynthetic fibers, macrosynthetic fibers or a
combination of both. Though it is effective instrengthening concrete, synthetic fibers are more
expensive compare to other alternative fiber
reinforcement. Studies on natural fibers such as
coconut fiber, jute fiber, human hair and etc. as fiber
reinforcement to concrete have been conducted bydifferent researchers. Pig Hair Fibers (PHF) are also
used as fiber reinforcement to concrete. Study showsthat using PHF as fiber reinforcement not only
improves the tensile strength of concrete but also its
compressive strength [6].
Upon discussing the potential feasibility of thetwo materials (GMS and PHF) as an alternative
component for concrete, it is quite interesting tostudy these materials when applied simultaneously
to concrete. This study could determine whether
their combination will yield better results or theirinteraction can have a negative effect on eachother’s characteristics. With the use of various
statistical analysis techniques, effect of interaction ofthese two materials on a particular response can beanalyzed. Statistical analysis like ANOVA candetermine the level of significance of each factor
and their interaction to the response. Along with
Response Surface Methodology (RSM), it can
provide non-linear analysis of results that captures
curvature on the response plot. RSM also enables theresearcher to determine optimum level of factors tocome up with the maximum possible yield and also
generate a prediction model for the fresh andhardened properties of concrete based on the
proportion of parameters involved.

Philippine shores specifically in the Manila Bay. To
extract its lime content, the shells were heated in apan while continuously stirring for three hours until
it became brittle. After heating, the shells werecrushed into powdered form. To ensure uniformity
in the particle size, crushed shells were sieve with 1
mm diameter opening. These powdered GMS wereused as partial cement substitute. PHF content and
GMS partial cement replacement are the main
parameters for this study.
Preparation of Materials
Collection and Processing

Properties of Materials
Physical Properties of each Materials

Mix Design Proportioning andPreparation
Design of Experiment Using CCD

Testing of Specimens
Fresh and Hardened Properties

Statistical Analysis
ANOVA, RSM and Cost Analysis
Fig. 1 Systematic plan of the study
Design of Experiment
The authors established proportions based on thebest mix design suggested by the previous studies of
Talagtag et al. [5] and Lejano et al. [6]. For PHF, 3
levels (0.6%, 0.8% & 1.0%) per volume of concreteof fibers were used. Also, 3 levels (5%, 10% &15%) of partial cement replacement were used for
GMS. These combinations were investigated for
different water-cement ratio of (0.4, 0.5 & 0.6).
Combination of these parameters is set using centralcomposite design (CCD) as shown in Fig. 2. A totalof 10 samples were made for each mix design forcompressive and tensile strength test.

METHODOLOGY
The systematic plan in Fig. 1 shows themethodology used in conducting the study.
Preparation and Processing of PHF and GMS
As discussed in the introduction, two wastematerials were used for this study. PHF and GMS
were processed and tested for their physicalproperties. PHF were collected from the Monterey
plant in Cavite, the fibers were washed and dried
under the sun for 48 hours to remove impurities
accumulated during the dehairing process. The GMS
used in this study are called the “Asian Green
Mussels”, there are the ones commonly found in

Making of Specimen and Testing
Mixing of concrete specimen was based on
ASTM C192 [7] or the Standard Method of Making
and Curing Test Specimen in the Laboratory. Dry
mixing technique was used for the mixing of
concrete. This mixing technique prevents
403

GEOMATE-Bangkok, Nov. 14-16, 2016

Note: Mix Design are coded as PHF percentage –GMS cement replacement – water cement ratio.

accumulation of air pockets in the mixture thus
making it more ideal when working with fiber
reinforced concrete.

Workability of Fresh Concrete
Shown in Table 1 are the results of the slump testfor each mix. Overall, the workability attained
during the experiment ranged from 10 to 38 mm.
Based on this result, PHF-GMS concrete would besuitable for foundation application with lightreinforcement. Furthermore, with high levels ofwater cement ratio, reduction of slump due to theincrease of fiber content tends to become lesscompared to samples with lower w/c ratio. Withhigher water cement ratio, this concrete mixture
could be used also for beams and columns. Fig. 3
shows the relation between the w/c ratio and therecorded slump of PHF-GMS concrete mix.

Fig.2 Central composite design of the study
The concrete were then molded into concrete
cylinders of 150 mm diameter and 300 mm in heightas test specimens. Each specimen was cured for 28
days before being tested using the Universal Testing
Machine (UTM) for its compressive and tensile
strength.

Average Slump vs. W/C Ratio

60

Slump (mm)RESULTS AND DISCUSSION
The results of the 20 mix design tested forworkability, compressive and split-tensile strength
are shown in Table 1.

0.6-5-0.4
1.0-5-0.4
0.6-15-0.4
1.0-15-0.4
0.6-5-0.6
1.0-5-0.6
0.6-15-0.6
1.0-15-0.6
0.46-10-0.5
1.14-10-0.5
0.8-1.59-0.5
0.8-18.41-0.5
0.8-10-0.33
0.8-10-0.67
0.8-10-0.5
0.8-10-0.5
0.8-10-0.5
0.8-10-0.5
0.8-10-0.5
0.8-10-0.5

Slump
(mm)
23
18
28
14
35
28
32
27
26
19
22
23
10
38
26
26
24
29
29
27

Compressive
Strength
(MPa)
32.23
31.27
25.81
23.29
15.13
16.19
14.31
11.21
25.41
20.00
20.14
15.92
31.21
13.36
26.34
26.03
25.71
24.00
23.96
23.22

43
38

36

40

Control

30.5
25.1

30

20.75

20
10
10
0

Table 1 Experiment results
Mix Design

PHF-GMS

49
50

0.3

TensileStrength
(MPa)
2.73
3.44
2.49
2.25
1.58
2.09
1.62
1.83
2.53
2.51
2.22
2.03
2.84
1.24
2.69
2.87
2.78
2.75
2.72
2.58

0.4

0.5
W/C Ratio

0.6

0.7

Fig. 3 Average slump for different w/c ratio
Comparing the slump of PHF-GMS concrete to
the control specimen, addition of PHF and GMS
resulted to a decrease in workability of concrete.
These results were expected due to the fact that
previous studies conducted on the application ofeach material also both resulted on a decrease inworkability.
Compressive Strength of ConcreteCompressive strength of each mix design wascomputed using the average of 5 concrete cylinders
tested. Like normal concrete, PHF-GMS concretealso experienced an inverse proportional relationshipbetween the w/c ratio and compressive strength asshown in Figure 4. Comparing the results of the
PHF-GMS combination to the control specimens, it
can be seen that the compressive strength of
concrete increased by 21.47% at 0.4 w/c, 32.71% at
0.5 w/c and 5.41% at 0.6 w/c. PHF-GMS concretewith W/C ratio of 0.5 yielded the highest
compressive strength gain, this is due of the fact thatthe optimum content for the two materials (0.8% and
10% respectively) were at the center of the designthat is where the 0.5 w/c ratio is also located.
However, among the three factors considered in this404

GEOMATE-Bangkok, Nov. 14-16, 2016

study, the w/c ratio still had the most effect to theresulting compressive strength as well as theworkability of concrete.
35.00

Maximum Compressive Strength per W/C ratio
31.27

30.00
Compressive Strength (MPa)PHF content and GMS partial cement substitute, thelatter had more significant effect on the compressive
strength of concrete. The fact that the cement
content of ordinary concrete greatly affects its
resulting properties, replacing some parts of it withpure lime from GMS would also have a significant
effect on strength of concrete.
After determining significant factors andinteraction, results were plotted to generate a
response surface that represents the behavior of thecompressive strength with respect to the PHF
reinforcement content and GMS partial cement
substitute. Combination of PHF and GMS areplotted for different water cement ratio as shown in
Figures 6 to 8.

32.23
26.53

25.00

PHF-GMS

26.34

Control
Specimen

19.84

20.00

16.19
13.36

15.00

15.36

10.00
5.00
0.00
0.3

0.4

0.5
W/C Ratio

0.6

0.7

Fig. 4 Maximum compressive strength at different
w/c ratio

Compressive Strength (MPa)Aside from the w/c ratio, another factor thatshowed significant effect to the compressive
strength is the GMS content. The graph shown in
Figure 5 describes the behavior of the compressive
strength with respect to the different amount of
GMS as partial cement substitute used in the mixture.
Results in Fig. 5 were analyzed using a fixed PHF
content of 0.8% per volume of concrete and 0.5 W/Cratio.
It can be seen that higher levels of GMS cement
replacement had a negative effect on the strength of
concrete. This decrease in strength may be
accounted to the unutilized lime-silica reaction ofexcess lime from higher GMS content to the silica
present in Portland-pozzolanic cement used.

Fig.6 Compressive strength 3D surface for 0.4 w/cCompressive Strength of Concrete vs
GMS content
26.34
27.00
25.00
23.00
21.00
19.00
17.00
15.00

Fig. 7 Compressive strength 3D surface for 0.5 w/c

20.14
15.92
1.59

10

18.41

GMS content (% Cement Replacement)

Fig. 5 Compressive strength for varying GMS
content
Fig. 8 Compressive strength 3D surface for 0.6 w/c

As for the results of the analysis of variance forcompressive strength, the 3 factors (PHF content,
GMS partial cement substitute and W/C ratio) were
all significant to the compressive strength of
concrete. Among these factors, w/c ratio has thehighest level of significance. This is due to the factthat W/C ratio had the most effect in terms ofincreasing/decreasing the workability of concrete as
it changes. Between the two materials incorporated,

Based on the 3d response surface plots forcompressive strength on different water cement ratio,
it was clearly observed that there is a nonlinear
behavior of compressive strength with the variation
GMS content from 5% to 15% for the 3 W/C ratio
level. At W/C ratio of 0.4, the maximum value ofcompressive strength was located near the minimum405

GEOMATE-Bangkok, Nov. 14-16, 2016

amount of GMS content considered which is at 5%.
However, as the W/C ratio increases from 0.5 and0.6 level, a shift on the optimum GMS content near
the center of surface was observed. This means that,high w/c ratio promotes better utilization of lime
content from GMS that is beneficial in the strength
gain of concrete with GMS as partial cement
substitute. Substituting type 1P (portland-pozzolan)
cement with GMS led to adequate lime-silica
reaction between the reactive silica of fly ash andlime extracted from the green mussel shells.

Generally, the tensile strength recorded follows
the same behavior as the compressive strength. It
can be seen also from the results that an increase intensile strength of concrete was experienced whenincreasing the fiber content to 1% compared to
specimen containing only a minimum amount of
0.6%. However, maintaining the PHF content of
concrete to its optimum percentage as suggested by
the previous study had the most positive effect interms of effectively increasing the tensile strength
while keeping the slump of concrete at an acceptablelevel. This behavior of the tensile strength of
concrete to varying PHF content in concrete wasillustrated in Fig. 9.

Tensile Strength (MPa)(2)

S = 6.402 – 15.660A – 0.019B + 63.038C(3)

Table 2 Optimum mix design

2.87

Factors/Response

2.51
2.20

0.8

TS = -2.299 + 5.808A + 0.101B + 13.118C– 0.156AB + 1.563AC + 0.303BC– 2.750A2 – 0.0078B2 – 22.138C2

The main objective of the optimization was togenerate the maximum possible compressive
strength while keeping the workability of the
mixture at an acceptable level, which for this case, aminimum slump of 25 mm. The optimum mix
design was determined using the numerical
optimization tool of Design Experts® as shown in
Table 2. These results showed that combining PHF
and GMS into concrete could yield a maximum
compressive strength of 27.394 MPa with anacceptable slump of 25 mm using the optimum
combination generated.

Tensile Strength of Concrete vs PHF
content

0.46(1)

where:CS = Compressive strength of concrete, MPa
TS = Split-Tensile strength of concrete, MPa
S = Slump of concrete, mm
A = Pig-hair fiber content, %B = Green mussel shell partial cement subs., %C = Water/cement ratio

Split Tensile Strength of Concrete

3.00
2.80
2.60
2.40
2.20
2.00

CS = 32.179 + 24.565A + 0.964B – 8.493C– 0.715AB + 9.000AC + 2.150BC– 17.042A2 – 0.093B2 – 83.018C2

PHF Content
GMS Content
W/C ratio
CS
Slump

1.14

PHF Content (%)

Optimum
Value
0.70
7.81
0.47
27.40
25

Unit
%
%
MPa
mm

Cost Analysis
Fig. 9 Tensile strength for varying PHF content

The cost analysis was focused on determining the
economy of the application of the two materials
under study. Direct cost such as materials, labor andmiscellaneous expenses were considered for PHF
and GMS. The fact that both of these materials wereconsidered as wastes, its materials cost could beassumed as zero. However, gathering and processingthese materials into concrete component induced
labor and operations cost. Table 3 shows the total
cost derived for each component material used in theconcrete. Also, shown in Table 4 is the costcomparison of ordinary concrete (OC) and PHFGMS concrete (PGC). Concrete strengths of control
specimen at different w/c ratio considered (15.36
MPa, 19.84 MPa and 26.53 MPa) were used as thebase strength for OC and PGC.

A 2.87 MPa tensile strength was recorded using
the optimum condition suggested by previous
studies that is about 74% higher compared to thecontrol specimen. It is evident that increasing thefiber content of concrete really increases its tensile
strength, however, further increase in PHF content
may result to reduced workability and accumulationof air voids that could cause strength reduction in
concrete.
Optimization
The regression models in Eqs. (1), (2) and (3)were used in predicting the optimum mix design for
the PHF and GMS combination.
406

GEOMATE-Bangkok, Nov. 14-16, 2016

Table 3 Unit price of materials
Material
PHF
GMS
CementGravelSand

Unit Cost
4.50
8.06
6.00
0.71
0.35

Using the mix design obtained from optimizationon actual specimens, PHF-GMS concrete obtained acompressive strength of 26.58 MPa and slump of
26mm. Results suggest that PHF-GMS could be
used in structures not requiring above 28 MPacompressive strength or high strength concrete.
These includes residential houses, low rise buildings
and also concrete pavements. And with the use of
GMS as a partial cement substitute, it could reduce
the overall cement requirement for a project thus
incurring savings and most importantly promotes theuse of environment friendly materials.

Php / kgPhp / kgPhp / kgPhp / kg
Php/kgComparing the results showed that incorporating
PHF and GMS reduces the cement content
requirement in concrete by the increase in w/c ratio
to achieve a particular compressive strength. Savingsof about 0.78%, 7.22% and 10.11% were recorded
compared to ordinary concrete with 0.6, 0.5 and 0.4
w/c ratio respectively.

ACKNOWLEDGMENT
Special thanks to the Department of Science andTechnology –Engineering
Research andDevelopment for Technology (DOST – ERDT) for
providing the financial support for this study.

Table 4 Cost comparison per strength requirement
CS
(MPa)
15.36
MPa
19.84
MPa
26.53
MPa

OC
PGC
OC
PGC
OC
PGC

W/C

GMS

PHF

0.60
0.63
0.50
0.58
0.40
0.48

0
9.86
0
9.22
0
8.3

0
0.68
0
0.68
0
0.70

Cost percu.m(Php)
2,815.43
2,793.43
3,166.70
2,938.08
3,693.61
3,320.16

REFERENCES[1] Mehta PK and Montiero PJM, “The futurechallenges in concrete technology”, Concrete:
Microstructure, Properties and Materials, 3rd ed.
2014, pp 636.
[2] Rubenstein M, “Emissions from the cement
industry”, The Global Network for Climate
Solutions (GNCS Factsheets), The Earth
Institute Columbia University, Retrieved fromhttp://www.theGNCS.org, 2012.
[3] Etuk BR, Etuk IF and Asuquo LO, “Feasibilityof using sea shells ash as admixture to concrete”,
J. of Environmental Science and Engineering,Vol. 1, 2012, pp. 121 - 127.
[4] Amarakoon GGNN, Koreeda T and Kawasaki S,“Effect on unconfined compressive strength ofsand test pieces cemented with calcium
phosphate compound”, Int. J. of GEOMATE,
Vol. 7, Dec. 2014, pp. 1070 - 1075.
[5] Talagtag R, Sarao III E, Ngo K and Lejano
B, “Utilization of perna viridis as partial
substitute to cement in concrete mix”,
Proceedings of the 2014 Seminar Workshop on
Utilization of Waste Materials, Philippines,
Philippines, Sep. 5, 2014 (In CD).
[6] Lejano B, Templonuevo J, Young CJ and
Zotomayor RL, “Properties of fresh andhardened concrete reinforced with hair from sus
scrofa domestica”, Proceedings of the SeminarWorkshop on Utilization of Waste Materials
2013, Philippines, Sep. 5-6, 2013 (In CD).
[7] ASTM International, “ASTM C192, Standard
Practice for Making and Curing Test Specimens
in the Laboratory”, http://www.astm.org, WestConshohocken, PA, 2016.

CONCLUSION
This study showed that the parametersconsidered pig-hair fiber reinforcement, GMS partial
cement substitute and water cement ratio had asignificant individual effect to the resulting
properties of concrete. Based on the results,
combining pig-hair fibers and GMS into concrete
further increased the compressive strength of
concrete when compared to the results obtained on
the research about application of GMS only by about
13.19% (from 23.27 MPa to 26.34 MPa) at 0.5 w/cratio. However, reduction in workability was
experienced upon the combination of these twomaterials into concrete. Generally, increasing theamount of PHF on the mixture tends to decrease theworkability of the fresh concrete. Adjustment of theoverall water content on the mix could be done toaddress the decrease in workability caused by fiber
reinforcement. Compressive strength as high as
32.23 MPa was recorded for PHF-GMS concrete at
0.4 w/c ratio which yielded a slump of only 14 mm
thus making it unacceptable for industry application.
The maximum compressive strength with above
minimum acceptable slump of 26mm was obtained
from
the
0.8%-10%-0.5(PHF-GMS-W/C)
combination that yielded 26.34 MPa which was
32.74% higher than ordinary concrete strength withthe same w/c ratio.
407

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

PULL-OUT STRENGTH OF AN EXPANSION STUD ANCHOR IN
CARBON FIBER REINFORCED CONCRETE
Gilford B. Estores1 and Bernardo A. Lejano2
Faculty, Mapua Institute of Technology, Philippines; 2Faculty, De La Salle University, Philippines

1

ABSTRACT
Carbon Fiber Reinforced Concrete (CFRC) is considered as an innovative structural material because of its
better performance characteristics when compared to conventional concrete. Its common applications where
expansion stud anchor connection is possible are in slabs on grade, wall panel, curtain walls, and pre-castelements. The design of this stud anchor in CFRC is of great interest to many structural engineers, however, nomodel is available as basis for its design. To develop such model, this study investigated the pull-out strength ofan expansion stud anchor embedded in CFRC as influenced by fiber volume content (Vf), fiber length (Lf),compressive strength and tensile strength. Three compression, three tension, and five pull-out CFRC specimens
each of different Vf (0.10%, 0.15%, 0.20%, 0.25%, 0.30%) and different Lf (19mm, 30mm, 38mm) wereprepared, tested, analyzed, and compared to control concrete specimens at design compressive strength of
21MPa. Tests results show that pull-out strength of an expansion stud anchor in CFRC is maximum at Vf =0.10% and Lf = 38mm. Among the parameters considered, tensile strength is the most significant contributing
factor that could influence the pull-out strength of stud anchor in CFRC. This is further verified numerically by a
FEM model with good agreement to the observed tensile strength. Finally, a Response Surface Methodology(RSM) model is proposed to predict the pull-out strength of an expansion stud anchor embedded in CFRC as
influenced by the fiber volume content, fiber length, compressive strength, and tensile strength.
Keywords: Pull-out Strength, Expansion Stud Anchor, Tensile Strength, Fiber Length, Fiber Volume Content

the maximum size of the coarse aggregate,
preferably at least twice as long as the maximum
size of coarse aggregate at minimum of 0.1% fiber
volume content to effectively bind coarse aggregate
to achieve a significant result for strengthening a
concrete structure.
One possible application of CFRC is in theanchorage connection of structural elements.
Assuming that the anchor bolts are designed
adequately, the pull-out strength of these anchors inconcrete is controlled by the failure mode of the baseconcrete, which could be concrete breakout, concrete
splitting, or frictional pull-out. Hence, the behaviorof the base concrete where these anchors will beembedded should be carefully considered in design.
It is believed that the base concrete will performbetter if it is reinforced with fibers. Fiber-reinforced
concrete (FRC) has been observed to perform better
compared to plain concrete. One study investigatedthe performance of adhesive anchor bolts inpolypropylene fiber reinforced concrete and in steel
fiber reinforced concrete [2]. It has been reported
that there was a significant increase in pullout load
capacity of adhesive anchors both in polypropylene
fiber reinforced concrete and in steel fiber reinforced
concrete compared to plain concrete. Another study
on pullout performance of a single headed anchor insteel fiber reinforced concrete has been investigated[3]. It was found out that the anchor’s pullout

INTRODUCTION
Synthetic fibers as replacement of steel
reinforcement had recently become the focus ofmany researchers. According to them, these fibers
have the potential to be used as reinforcement in
concrete to improve certain physical properties.
Among the synthetic fibers, carbon has beenconcluded in many studies to have excellent physical,
mechanical and dynamic properties and can beutilized more effectively as reinforcement in
concrete material. It is believed that addition ofcarbon fibers to concrete effectively increases thestrength and toughness of concrete. The inclusion ofcarbon fiber to concrete is termed as CFRC for
carbon fiber reinforced concrete. CFRC is proven tohave high resistance to corrosion which makes it
more durable. It has been successfully applied tomany civil engineering projects such as impact
resisting structures, precast elements, panels, bridge
deck, slabs-on-grade, pavements and curtain walls. It
has been claimed by some researchers that chopped
carbon fibers, when included within concrete withthe appropriate fiber length and volume fraction, canmodify the tensile strength, flexural strength,
toughness, impact resistance, and fracture energy of
the concrete [1]. They recommended that the
chopped carbon fibers regardless of its type (PAN or
Pitch) should have an average length of not less than408

GEOMATE- Bangkok, Nov. 14-16, 2016

capacity increased with the addition of steel fibers toconcrete.
In this study, CFRC is investigated as basematerial of expansion anchor connection. As of now,there is no model that can be used as basis for its
design. The pull-out strength of an expansion anchor
in CFRC is expected to behave differently fromusual ordinary concrete. Specifically, this study
investigated the pull-out strength of a single
expansion stud anchor in CFRC considering theconcrete breakout failure mode as influenced byfiber volume content, fiber length, tensile strengthand compressive strength. This research will givesignificant information that is much needed bystructural engineers in designing an expansion studanchor in CFRC which will lead to the solution
involving the safety and the economic aspects of its
design. Moreover, this study also promotes theapplication of anchorage in CFRC as well as theutilization of carbon fiber as reinforcement in
concrete structures that would result in valuing andincreasing awareness of the carbon fiber as
innovative construction alternative material. Lastly,
the additional benefit of anchorage in CFRC is theease of drilling for post-installed anchor bolts.

Table 2

Properties of PAN-based HT carbon fibers

Lf
(mm)

Ft
(MPa)

E
(MPa)

Spec.
Gravity

Width
(mm)

4510

2310001.8

3.0

Notes: F t = Tensile Strength of Carbon Fibers
E = Modulus of Elasticity

METHODS

The concrete matrix was composed of cement(C), water (W), fine aggregates (FA), coarse
aggregates (CA), and superplasticizer (SP). A highearly strength and with improved workability
Portland cement that meets the ASTM standard
specification C595 [5] was used in this study. The
water used was clean and of good quality. Crushed
coarse aggregates with maximum size of 19mm
having a mass density of 1592 kg/m3 and 1.01%
absorption were used. White sand with mass densityof 1551 kg/m3, fineness modulus of 2.4 and having a
3.01% water absorption were used as fine aggregates.
A superplasticizer was added to ensure that fresh
CFRC mix is workable.

Materials

Specimens

The components of the specimens used in thisstudy were expansion stud anchors and thecomposite base materials. The composite basematerial is composed of the carbon fibers (CF) andthe concrete matrix. An expansion stud anchor usedin this study was a medium-duty anchor with 10 mmdiameter with a total length of 90 mm and with yieldstrength of 640 MPa. The technical data of thisanchor such as required torque (T), standard
effective embedment depth (h e ), drilling depth (h),drilled hole diameter (D h ), and minimum basethickness (H min ) is presented in Table 1.
The fiber type used in this investigation was
0.111 mm thick chopped PAN-based high tensile(HT) strength carbon fibers. According to ACI 544[4], the length of carbon fibers may vary from 5mm
to 50mm, but with the predominance of demand for
19mm or 38mm fiber length. The selected fiber
lengths used in this investigation were 19mm, 30mm,
and 38mm. The properties of the PAN-based HT
carbon fibers used are presented in Table 2.

The design mix of the composite base materials
were based on the compressive strength of 21MPa
considering the 25mm to 100mm slump, 2%
entrapped air, and water-cement ratio of 0.68. Threecompression, three tension, and five pull-out CFRC
specimens for each of the different fiber volume
contents, Vf (0.10%, 0.15%, 0.20%, 0.25%, 0.30%)
and of different fiber lengths, Lf (19mm, 30mm,38mm) were prepared, tested, and compared toconcrete without fiber as the control specimen. The
mix proportions of the control and CFRC specimens
are given in Table 3.

T
(Nm)

he
(mm)

h
(mm)

Dh
(mm)

H min
(mm)

45

60

80

10

120

Mix proportions of the specimens

Vf
(%)

W
(kg/
m3)

C
(kg/
m3)

CA
(kg/
m3)

FA
(kg/
m3)

CF
(kg/
m3)

SP
(gm/kg
of C)

0

182.
08
182.
08
182.
08
182.
08
182.
08

301.
47
301.
47
301.
47
301.
47
301.
47

1054
.71
1054
.71
1054
.71
1054
.71
1054
.71

841.
07
839.
52
838.
71
837.
93
837.
16

-

-

1.8

3

2.7

3

3.6

3

4.5

3

0.1

Table 1 Technical data for expansion stud anchor
Type

Table 3

0.15
0.2
0.25

409

GEOMATE- Bangkok, Nov. 14-16, 2016

0.3

182.
08

301.
47

1054
.71

836.
38

5.4

3

Before casting all the specimens, the workabilityof each mix was checked by measuring its slump as
per ASTM C143 [6].
The compressive strength specimens were tested
using 100mm x 200mm cylinders as per ASTM
C39-05 [7] after 28 days of curing period. The
tensile strength specimens using a dumbbell shapewith a critical section of 75mm x 50mm, and agauge length of 300mm were tested after 28 days ofcuring period. The test set-up for direct tensile
strength and its failure mode at the critical section
are shown in Fig. 1. The base material specimens inrectangular solid shape measuring 360mm x 360mm
x 150mm were made and cured for 28 days. Then,the stud anchors were set in these base materials
following the setting instruction recommended by
the manufacturer and tested them for pull-out
strength in accordance to ASTM E 488-96 [8]. The
set-up of pull-out strength test for expansion studanchors and its concrete breakout failure mode arepresented in Fig. 2. The observed pull-out strengthof the stud anchor in concrete base material without
fiber was compared to ETAG 001 (Guideline forEuropean Approval of Metal Anchors) [9], and ACI
355.2 [10] and NSCP 2010 (National Structural
Code of the Philippines) [11] equations forverification. ETAG 001 equation for concrete
breakout of a single anchor in non-cracked plain
concrete is given by0
N Rk
f ck ,cube hef,c = 10.1

1.5

(2)

While, the NSCP 2010 and ACI 355.2 equation for
concrete breakout of a single post-installed anchor in
non-cracked plain concrete is given by

f 'c hef

1.5

Fig. 2Pull-out test set up and failure mode

In this study, FEM was applied to simulate, andverify the tensile stress response of the CFRC basematerial specimens subjected to tensile loading
applied into a single expansion stud anchor. Since
numerical analysis using FEM may lead to a verylarge equations and complex solutions, the use of
FEM software ABAQUS was used to verify andcompare the tensile stress response of CFRC basematerial specimens subjected to tensile loading
applied to a single expansion anchor against theactual tensile stress of the specimens. The basematerial is modelled as an isotropic 2D elastic
material under tensile load induced by an expansion
stud anchor as shown in Fig. 3 with assumed Poisson
ratio of 0.20.

The compressive strength using cube specimen wascomputed using the equation of Kumavat HR andPatel VJ [12],N b = 7(1.4)

Tension test set up and failure mode

Finite Element Modeling (FEM)(1)

f ck ,cube = f 'c / 0.95

Fig. 1(3)410

GEOMATE- Bangkok, Nov. 14-16, 2016

∧

∧

k

∧

y = β 0 + ∑ β i xi
i =1

(6)
∧

∧

k

∧

k

∧

∧

y = β 0 + ∑ β i xi + ∑ β ii xi2 + ∑ ∑ β ij xi x j
i =1

i =1

i< j

(7)
∧

respectively, where, β is the least squares estimateof model coefficients. The next step is the evaluationof the adequacy of the fitted model. This is to ensurethat the recommended model will give a satisfactory
estimate of the true system. The adjusted coefficientof multiple determination or R2 adj was used in thisstudy instead of R2 because in most cases, it doesn’t
always increase as the terms are added to the modelwith k regressors [14]. It is a measure that estimates
Pearson’s correlation ratio with value from 0 to 1
and is defined as:n − 1  SS E 
2
(8)


Radj
=1−
n − p  SST 
where, SS E is the sum of squares of the residuals,
SS T is the total sum of squares and p = k+1 degrees
of freedom. The fitted model was also subjected totest of significance by F-test at α = 0.05. The bestmodel is the one that has highest R2 adj and has
least error.
Error metric was defined to each model andcompared. The metric of comparison used aside
from R2 adj , is the mean square of error or MSE. The
error metric is given as

Fig. 3 FEM model
The experimental parameters used in thesimulation of tensile stress (FEM ft ) are compressive
strength (f c ), mass density of base material (w c ), andpull-out strength (N), while normal force through theanchor’s sleeve expansion (F exp ) was derived using
the equation,

N = uFexp(4)where, u is the coefficient of friction and wasassumed 0.372. This coefficient is nearly equal tothe value cited in CEB (Comite Euro-International
du Beton) [13], where u of expansion anchor varies
from 0.4 to 0.6. The modulus of elasticity, E c of thespecimens were computed using the equation given
by NSCP 2010,
Ec = wc

1.5

0.043

f 'c

(5)

∑  y − y 
∧

MSE =

Response Surface Methodology

2

(9)

nσ y2

∧

Response Surface Methodology (RSM) canmodel linear and non-linear dynamics, andstochastic phenomenon. The data requirement for agiven output is low and the errors are assumed to berandom. Supposed that Y is the response of interestand x 1, x 2, x 3 , x 4 are the predictor variables. The
response of interest in this study can be expressed as
Y = f(x 1, x 2, x 3 , x 4 ). The function f(x 1 , x 2 , x 3 , x 4 )denotes the response surface. The first step in thisprocess is to find an appropriate approximation for
the true relationships between the response and thepredictor variables [14]. In this study, the responseof interest is the pull-out strength of an expansion
stud anchor in CFRC, while the predictor variables
are the fiber volume content, fiber length, tensile
strength and compressive strength at standard h ef =
60mm. First-order and second-order RSM models
were considered in developing the proposed modelin this study. The fitted model of the first-order andsecond-order are generally defined aswhere, y is the observed value, y is the predictedvalue, and σ y2 is the variance of the observed value.
RESULTS AND DISCUSSIONS
Slumps of Fresh Concrete
The workability of fresh CFRC mix and plain
concrete resulted from the slump tests are presented
in Fig. 4. The slump of CFRC mixes range from75mm down to 47mm, while the control specimen
has a slump of 80mm. It reveals that all the mixes
have passed the slump requirement from 25mm to100mm. The addition of superplasticizer to CFRC
mixes would have caused this satisfaction in slump.
The result indicates that regardless of the fiber
length used, the slump tends to decrease consistently
with increasing fiber volume content from 0.1% to
0.3%.

411

GEOMATE- Bangkok, Nov. 14-16, 2016

specimen.
On the other hand, Fig. 6 and Fig. 7 consistently
shows a positive linear trend with the addition of Lf= 19mm and Lf = 38mm. With 19mm, it is evident
that pull-out strength is maximum at f c = 19.84MPa
and at f t = 1.903MPa. This pull-out measures 24.33
KN andFig. 4 Effect of Vf to slump of CFRC mix
Pull-out Strength of an Expansion Stud Anchor
Test data on pull-out strength of the expansion
stud anchor considering the concrete breakout failuremode of different base material specimens areshown in Fig. 5. The control specimen’s pull-out
strength was measured 21.53 KN. This result issimilar to 21.63 KN computed from Eq. (1). It isalso close to 20.46 KN calculated from Eq. (3).
Moreover, Fig. 5 shows that in 19mm fiber length
case, pull-out strength tends to increase when fiber
volume content increases. The pull-out strength
measures 21.52 KN initially at 0.1% fiber volume
content and increased to 24.33 KN at 0.30% fiber
volume by 13.03%.

Fig. 6 Influence of compressive stress to pull-out
strength of CFRC
is 13.03% more than the pull-out strength of thecontrol specimen at f c = 20.17MPa and at f t =1.683MPa. On the other hand, with 38mm, the pullout strength is highest at f c = 21.85 MPa and at f t =2.047MPa. This pull-out strength measures 26.82KN
and is 24.61% more than the pull-out strength of thecontrol specimen at f c = 20.17MPa and at f t =1.683MPa. It is also evident that compressive stressof CFRC at 19mm and 38mm slightly affect thepull-out strength, but significantly affect by thetensile stress. While, no trend is observed to pull-out
strength with the addition of Lf = 30mm both forcompressive stress and tensile stress.

Fig. 5 Influence of fiber volume content to pull-out
strength of CFRC
In the case of Lf = 30mm and Lf = 38mm, a
negative trend between the pull-out strength andfiber volume content is noticed. Despite this, it isobserved that there was a significant increased ofpull-out strength with the addtion of Lf = 38mm at
Vf = 0.10% and Vf = 0.15% by 24.61% and 11.83%
respectively. At Lf = 30mm, however, no significant
increased was observed. Among the cases, it has
found out that the maximum pull-out strength
occurred with the addition of Lf = 38mm at Vf =0.1% by 24.61% increase compared to control

Influence of tensile stress to pull-out
Fig. 7
strength of CFRC
FEM of CFRC Tensile Strength
The comparison between the observed tensile
strengths and the FEM tensile strengths resulted
412

GEOMATE- Bangkok, Nov. 14-16, 2016

from the simulation of the tensile strength responseof the CFRC base material specimens subjected totensile loading applied into a single expansion studanchor using the ABAQUS is presented in Fig. 8. It
can be observed that there is a good agreement
between the observed tensile stress values andsimulated tensile stress values subjected to a tensileloading applied into a single expansion stud anchor
embedded in both control and CFRC specimens with
R adj of 0.818 and 0.999997 respectively. This result
implies only that the tensile strength is the mostimportant influencing factor to predict the pull-out
strength of an expansion stud anchor embedded in
CFRC. This is similar from finding of a previous
study where the pullout load capacity of a headed
anchor was dominated by the tensile strength of its
composite base material [15].

volume content, compressive strength, and tensile
strength as its predictor variables which is given by

N = 0.038Lf − 8.785Vf + 0.146 f c + 8.237 f t + 5.7(10)
CONCLUSION
In general, it is concluded that CFRC increased
the pull-out strength of the expansion stud anchor.
Test results show that pull-out strength is highest at
Vf = 0.10% and Lf = 38mm. Moreover, among thepredictors considered, tensile strength turned out as
the most significant variable to influence the pull-out
strength of an expansion stud anchor in CFRC. Thisfinding is further verified by numerical analysis
using the FEM Software, ABAQUS. It is also
concluded that pull-out strength of an expansion studanchor in CFRC is best predicted by RSM modelwith fiber length, fiber volume content, compressive
strength, and tensile strength as its predictor
variables.
ACKNOWLEDGEMENTS
The carbon fiber used in this study was UT70-20
TORAYCA, and the expansion stud anchor used inthis study was HST M10X90/10 from Hilti. The
authors would like to thank DLSU, ERDT, Mapua
Institute of Technology, and Alphatec Chemical
Corp. for their valuable supports in this study.
REFERENCES[1] Tezuka M, Awata, M and Shiraki A, “Carbon
Fiber-Reinforced Concrete and Method for
Preparing the Same”, United States Patent,
Patent Number 5685902, 1997.
[2] Coventry K, Richardson A, McIntyre C andAresh B, “Pullout Performance of Chemical
Anchor Bolts in Fibre Concrete”, Fibre
Concrete, Prague, 2011.
[3] AlTaan SA, Mohammed AA and Al-Jaffal AA,“Breakout Capacity of Headed Anchors in Steel
Fibre Normal and High Strength Concrete”,
Asian Journal of Applied Sciences, Vol. 5 (7),2012, pp. 485-496.
[4] ACI Committee 544, “State-of-the-art Report on
Fiber Reinforced Concrete”, 2000, ch. 4.
[5] ASTM C595, “Standard Specification for
Blended Hydraulic Cements”, ASTM Standards
in Building Codes, Vol. 1 (43), 2006.
[6] ASTM C143, “Standard Test Method for Slumpof Hydraulic Cement Concrete”, ASTM
Standards in Building Codes, Vol. 1 (43), 2006.
[7] ASTM C39-05, “Standard Test Method forCompressive Strength of Cylindrical Concrete
Specimens”, ASTM Standards in Building
Codes, Vol. 1 (43), 2006.

Fig. 8 Comparison of observed f t and FEM f t
RSM Model
Table 4 Performance of RSM models
Model
N=f(f t )N=f(f c ,f t )N=f(Lf,Vf,f c )N=f(Lf,Vf,f t )N=f(Lf,Vf,f c ,f t )N=f(Lf,Vf)

Deg
1
1
1
1
1
2

R adj
0.82
0.81
0.60
0.85
0.85
0.76

MSE
0.29
0.28
0.48
0.21
0.19
0.26

Adequacy
Yes
Yes
Yes
Yes
Yes
Yes

Table 4 shows the performance of each adequate
RSM model to predict the pull-out strength of an
expansion stud anchor in CFRC as influence by
combination of different predictor variables such as
Lf, Vf, f c , and f t . Each combination was tested forthe adequacy of the first-order and second-order
RSM models using the R adj , MSE, and F-test at α =0.05. It is noticeable that tensile strength is the mostsignificant lone predictor with R adj of 0.82 and MSE
of 0.29. However, it is found out that the pull-out
strength of an expansion stud anchor in CFRC isbest predicted by RSM model with fiber length, fiber
413

GEOMATE- Bangkok, Nov. 14-16, 2016[8] ASTM E 488-96, “Standard Test Methods forStrength of Anchors in Concrete and Masonry
Elements”, ASTM Standards in Building Codes,
Vol. 4 (43), 2006.
[9] ETAG 001, “Guideline for European TechnicalApproval of Metal Anchors for Use inConcrete”, Annex C: Design Method for
Anchorages, Brussels, Belgium, 1991, pp. 16-21.
[10] ACI Committee 355.2, “Evaluating the
Performance of Post-Installed MechanicalAnchors in Concrete (ACI 355.2-00)”, Reported
by ACI Committee 355, 2000.
[11] NSCP 2010, “National Structural Code of thePhilippines 6th Ed.”, Anchorage to Concrete,Vol. 1, 2010, pp. 136-144.
[12] Kumavat HR and Patel VJ, “Factors Influencing
the Strength Relationship of Concrete Cube andStandard Cylinder”, International Journal of
Innovative
Technology
andExploringEngineering, Vol. 3 (8), 2014, pp. 76-79.
[13] Comite Euro-International du Beton (CEB),“Fastenings to Concrete and Masonry
Structures: State of the Art Report”, Thomas
Telford Services Ltd, London, UK, 1994, pp.
93-94.
[14] Montgomery DC and Runger GC, “Applied
Statistics and Probability for Engineers 5th ed.
SI. Version”, John Wiley and Sons, Inc.. Printed
in Asia, 2011, pp. 619-636.
[15] Qian S and Li VC, “Headed Anchor/Engineered
Cementitious Composites (ECC) Pullout
Behavior”, Journal of Advanced Concrete
Technology, Vol. 9 (3), 2011, pp. 339-351.

414

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EVALUATION ON THE EFFECT OF BAMBOO STRIPSREINFORCEMENT TO THE MECHANICAL AND PHYSICALPROPERTIES OF RICE HULL PARTICLE BOARD
Hermie M. del Pilar1, Jecelle Elyssa E. Calip 1, Aldwin Christopher A. Loyola1 and Neil Patrick C. Nase1
1
Department of Civil Engineering, Malayan Colleges Laguna,Pulo Diezmo Road Cabuyao City, Philippines

ABSTRACT
Adhering on use of sustainable materials and waste management, this study, focused on the effect ofbamboo strips reinforcement for rice hull particle board. The evaluation was based on the resulting physical andmechanical properties of the products. Both reinforced and unreinforced boards were made using 70:30, treatedpolystyrene: rice hull weight ratio with 5mm thickness. The reinforced boards were prepared with variation inorientation at 0O, 45O, 90O weaved bamboo strips. The boards were tested using Water Absorption (WA) andThickness Swelling (TS) for the physical property, while Modulus of Rupture (MOR), Internal Bonding Strength(IBS) and Impact Resistance (IR) were tested for the Mechanical Property. Each test subjected 12 specimens
consisting 3 for each of the 3 orientations of the reinforced boards, while the other 3 was for the unreinforced.
Results indicated that the 45O orientation gave the highest MOR performance at 40.22MPa significantly high,while the 90O demonstrates highest in IR test at 6.6 N-m, not significantly different with the other boards. The
unreinforced board yields 17.35 MPa for MOR but highest in the IBS test at 2.475 MPa, significantly higher
with other samples, and an IR of 5.30 N-m. In the WA and TS tests, the unreinforced board yield highest at
8.04% and 1.87% respectively, both not significantly different from the rest. With these results, it was evaluatedthat reinforcing the particle board with bamboo strips greatly enhances the mechanical and physical properties of
the average rice hull particle board.
Keywords: Rice Hull Particle Board, Bamboo Strips Reinforcement, Treated Polystyrene, Mechanical Property,
Physical Property

also be used as surface reinforcement for composite
boards [5].

INTRODUCTION
Globally, human population increases andso is the demand for construction materials. There isa need to the sustainability of construction materials
that finding alternatives is a must.

Another material being considered as asustainable substitute for wood and wood-based
board products is the rice husk. Rice husk isabundant in tropical countries such as thePhilippines [6]. The abundance of rice husk, its low
bulk density (90-150kg/m3), toughness, abrasive
property ,resistance to weathering and uniquecomposition are the characteristics why it can beimportant to the construction industry (Johnson &Bin Yinus, 2009).

Bamboo has proven its worth inconstruction due to its inherent strength anddurability [1]. Tropical countries have adequate
supply of bamboo [2] and because it is renewable
and can be easily harvested, it is considered as asustainable
construction
material.
Othercharacteristics such as being light, flexible, andbiodegradable add to bamboo’s advantage. It canalso be used as a reinforcement material when
extracted to appropriate fiber forms for a desired
purpose [3]. Bamboo can be a substitute for wood as
construction material because their characteristics
are comparable if not greater in some respect thanwood [2]. This material has double tensile strengththan lumber while the compressive strength isestimated to be 1.5 times greater [4]. Also, bamboo
has a long line of household products in whichbamboo mat is one of them. The bamboo mat canIn the study of Lagrimas et. al. (2014), thepolystyrene to rice hull mix ratio of 70:30 gave themost ideal properties in terms of Modulus ofRapture (MOR), Modulus Of Elasticity (MOE),
Water Absorption (WA) and Thickness Swelling(TS) compared to other mix ratios such as 80:20 and
60:40 [7]. However, particleboards made of ricehull demonstrate low mechanical property like the
MOR as compared to other particleboards. The
common MOR value for a particle board records an

415

GEOMATE- Bangkok, Nov. 14-16, 2016average of 16.6 MPa, while only 14.53 MPa for arice hull particle board [5].
This study aims to evaluate the effect ofbamboo reinforcement to the mechanical properties
and physical properties of rice hull particleboard
using the bamboo strips.

Reinforcements
Bamboo of the type kawayang tinik(Bambusa blumeana) were purchased from the local
bamboo dealer of Bgy. Dita, National Highway Sta.
Rosa City. The inner part of the bamboo is
processed into thin strips with dimensions of 200
mm x 4 mm x 1 mm. Bamboo strips were weaved
into mats with three different orientations namely 0o,45o, and 90o and these were equally spaced at 12.50
mm. The arranged bamboo strips are then brushedwith wood glue to maintain its position andorientation.

The properties of rice hull particle board
reinforced with bamboo strips arranged withdifferent orientations were evaluated. The
mechanical properties: modulus of rupture (MOR);
internal bond strength (IB); impact strength (IS) andphysical properties: density; water absorption (WA)and thickness swelling (TS) were measured. The
properties of the reinforced particle board werecompared to the unreinforced rice hull particle board.
Evaluation of results was done through Analysis of
Variance (95% confidence level). The inferential
statistic was used to identify if there is a significant
difference on the properties of the reinforced andunreinforced rice hull particle boards. This study
also helps in identifying the ideal orientation of thebamboo reinforcement for rice hull particleboards.

Fig. 1 Different Orientations of Bamboo Strips

Production of Rice Hull Particle Board

METHODOLOGY

Rice hull particle boards were preparedwith the use of two roll mill and a hydraulic press
found at the Industrial Technology Institute (ITDI)
of the Department of Science and Technology(DOST), located at Bicutan, Taguig City,
Philippines. The two roll mill was set to atemperature of 130oC and used in mixing the binderand rice hull. The ratio of the mixture was 70:30
binder: rice hull weight ratio. The binder-rice hull
mixture was poured constantly in the rotating tworoll mill to spread and mixed evenly, and this should
be repeated until there is a visual homogenous
mixture. The cake like mixture were then placed in a
200 mm x 200 mm x 5 mm to produce theunreinforced particle board by compressing and 200
mm x 200 mm x 2 mm for the production of thereinforced particle board. The molders with mixture
will be subject to the hydraulic press with atemperature of 180oC for about 5 minutes and thensubjected for another 5 minute cold press.

The proceeding sections discuss how thematerials were sourced and prepared, how theparticle boards were produced and the manner on
how the board specimens were tested.
Materials
The required materials to produce the particle
boards are described and prepared as follows.
Rice Hull
Rice hull was collected from Queen
Elizabeth Rice Mill Honorville, Macabling Sta Rosa
City Laguna with moisture content ranging between
2 to 5 percent. Rice hull were sieved using the sieve
number 12 and pan. Only the retained rice hull on
the sieve number 12, with a diameter of 1.68 mm,
was used.
BinderPolystyrene wastes were gathered fromArmstrong Trading Malitlit Road Bgy. Malitlit, Sta.
Rosa Laguna and dissolved in a petroleum naphthawith a weight ratio of 1:2 respectively. The weightratio was determined through several trials in which
the treated polystyrene was liquefied until it reached
its optimum paste – like state. Pure polystyrene
waste can be melted directly to bind the board
particles, but this will not produce good binding
effect.

Three types of reinforced particle boards
were made and each contained different orientationof the bamboo strips namely 0O, 45O and 90O.
Bamboo strip reinforcement were sandwiched
between the two 200 mm x 200 mm x 2 mm size
particle board. The composite boards were placed in
a 200 mm x 200 mm x 5 mm molder and then it wasinserted in the hydraulic press again with the same
temperature. Heat press was applied for about 10
minutes and another10 minutes for the cold press.
A total of 8 boards were prepared, 2 for theunreinforced and 6 for the reinforced. There were 3416

GEOMATE- Bangkok, Nov. 14-16, 2016

types of reinforced boards, and each type requires 2
boards, totaling to 6 reinforced boards.
Test Procedures
In each type of board: unreinforced, 0O
reinforced, 45O reinforced and 90O reinforced, thefollowing quantity and dimension of test specimens
were drawn from 2 boards (total of 8), as shown by
the table below.

(A)

Fig.2: (A) MOR test set-up; (B) IBS test set-up

Table 1: Quantity and Dimension of Test Specimens
for each type of board
Type of Test

Dimension (mm)

MOR
IBS
IR
WA & TS

200mm x 50mm
50mm x 50mm
50mm x 10mm
50mm x 50mm

(B)

Impact Resistance (IR)
Following ASTM E23, a board size of 10
mm × 50 mm was used as specimen for the Charpy
impact test. The sample was set horizontally fromthe base knocked by a hammer in a swinging
pendulum manner. With this test, the impact
resistance or the energy absorbed per unit thickness
was determined.

Quantity3333Modulus of Rupture (MOR)
The three point flexural test was performed
based on ASTM D1037-12 to measure the modulus
of rupture of the rice hull particleboard. The
Shimadzu Universal Testing Machine was used
where the test samples were placed longitudinally on
two tension rods with measured distances. The third
tension rod was lowered slowly and driven to the
mid of specimen.
Fig. 3: Impact resistance test using the Humburg
Pendulum Impact Test

The specimens were subjected to a threepoint strain and break when the maximum load was
attained. Modulus of rupture can be determinedusing the formula:
3PL
MOR =
(1)
2

Water Absorption

2bd

Based on ASTM D1037-99, the test
specimens with dimensions 50 mm x 50 mm were
initially weighed. Specimens were then submergedin distilled water for the next 24 hours then removed
and re-weighed. Percent water absorption wasdetermined using the formula.

where P is the maximum applied load, L is length of
the support, b is the base and d is the depth of thetest specimen tested.
Internal Bond Strength (IBS)% Water absorption =In accordance to ASTM D1037, Standard Test
Method for Evaluating Properties of Wood-Base Fiber andParticle Panel Materials, test specimens must have a 50
mm by 50 mm base with a height of 5 mm. The test
specimens were fastened to the grip fixtures of themachine which induces load to test subject’s surface until
it reaches failure. Internal bonding strength can be
described by the formula:Thickness Swelling

Wet Mass − Dry Mass
Dry Mass

x 100% (3)In reference to ASTM D1037-99, the 50
mm x 50 mm specimens used for WA test can also
be used to measure the thickness swelling. The
initial and final thicknesses were measured todetermine the change in thickness. The formula for
thickness growth (Gt) is given by:

F

IBS =
(2)
A
where F is the maximum load at failure and A is the
cross sectional area of the test specimen.

Gt =

417

Final Thickness − Initial Thickness
x 100% (4)Initial Thickness

GEOMATE- Bangkok, Nov. 14-16, 2016

Statistical Tool

Effect of Bamboo Reinforcement on
MOR (MPa)

The result for each test were statisticallyanalyzed using the Analysis of Variance (ANOVA)
with 95% confidence level. The tests for thesignificant difference between the means of theproperties of reinforced particle boards of different
orientations (0O, 45O and 90O) were evaluated. Also,t – test analysis was used to determine if there is asignificant difference between the mean of
unreinforced against mean of the least performing
reinforced particle board.

MOR

60

20

40.22

38.85

0 DEG
45 DEG
Reinfocement

90 DEG

34.6

40
17.35

0
URPB

Fig.4: Effect of bamboo strips reinforcement
RESULTS AND DISCUSSION

The physical properties, (WA & TS) of 90O
reinforcement in Table 2 show best results comparedto other type of boards, though all exhibit low
percentage of absorption and swelling. The least
performing in terms of physical properties is theunreinforced board. It is also clear that the water
absorption and thickness swelling are correlated asshown in Fig. 5.

The mean of the physical and mechanicalproperties of the particle boards are shown in Table
2 below.
Table 2: Mean of Mechanical and Physical
Properties of Particle Board
TYPE

Unreinforced

Reinforced

0o
45o
90o

ρ

kg/m3

Mechanical Properties

Physical
Properties

Water Absorption and Thickness Swelling

MOR
MPa

IBS
MPa

IR
Nm

WA
(%)

TS
(%)

10.00

0.925

17.35
(2.44)

2.48(0.20)

5.30(0.10)

8.04(4.53)

1.87(1.01)

8.00

1120

34.60
(1.17)

2.07(0.22)

5.77(0.38)

3.94(4.33)

1.50(0.02)

1130

40.22(1.17)

1.59(0.23)

6.08(0.38)

2.50(3.22)

0.74(0.97)

2.00

1120

38.85
(2.10)

1.91(0.02)

6.60(0.17)

2.21(0.99)

0.00(0.00)

0.00

Numbers in parenthesis are the standard deviation

6.00

WA

4.00

TS

URPB

0 DEG 45 DEG 90 DEG

Fig. 5: Water absorption and thickness swelling
relation

Table 2 shows that the internal bondstrength (IBS) of the unreinforced particle board(URPB) has the highest bonding strength whencompared to the reinforced particle boards (RPBs).
This indicates that the particles’ bonding is affectedby the reinforcement. This could be explained by thefact that the bamboo strips reinforcement has
decreased the contact area of the particles, where thetwo (2mm) particle boards sandwiched thereinforcement. The MOR and IR however, are muchgreater for the reinforced than the unreinforced. The
results of 45o and 90o orientation of the RPB
governed in the MOR and IR respectively. This
result also shows that the MOR and IR were affectedby the density, the orientation of the reinforcement,
and the quantity of bamboo reinforcement.

The table below, Table 3 summarizes theresults in determining if there is a significant
difference on the properties of the following types ofboards using the One Way Analysis of Variance forthe 3 reinforced boards, while the statistical t- test
was used to compare the least means of reinforced
boards against the means of the unreinforced boards.
Table 3: Test for the difference in properties of theparticle boards
Comparison
between
3 Types ofreinforced
Rice HullParticle
Board
Unreinforced
andReinforced

Figure 4 shows the effect of bamboo strips
reinforcement on the MOR of the rice hull particle
board.

418

MOR
(MPa)

IBS
(MPa)

IR
(N-m)

Physical
Properties
WATS(%) (%)

S

S

NS

NS

S

S

S

NS

NS

NS

Mechanical Properties

GEOMATE- Bangkok, Nov. 14-16, 2016

S: Significant;

NS: Not Significant
4. The use of bamboo strips reinforcement does
not affect significantly the water absorption andthe thickness swelling, though the reinforcement
can increase the water proofing property of theboards. Also, result shows that a 90O orientation
for reinforcement is ideal since it has the least
WA and TS.

The result shows that the Modulus of Rupture,
Internal Bond Strength and Thickness Swelling weresignificantly different among the 3 types ofreinforcements of rice hull particle boards. The
impact resistance and water absorption however
gives no significant difference on the results of theirmean.

ACKNOWLEDGEMENTS
In comparing means of the least value of reinforced
particle boards against the means of the unreinforced
board, the properties which show significant
difference are the Modulus of Rupture and theInternal Bond Strength. All other properties show no
significant difference.

The authors would like to thank the Industrial
Technology Development Institute (ITDI) of the
Department of Science and Technology (DOST) in
Taguig City and the Malayan Colleges Laguna(MCL) in Cabuyao Laguna, Philippines for allowing
the use of their equipment for the testing in researchstudy.

CONCLUSIONREFERENCES
Based on the results on mechanical and physicalproperties of the reinforced rice hull particle board,the following conclusions were drawn.

[1] Syeda, A., & Shrujal, B. (2014). A Case Studyon Bamboo as Green Building. International
Journal of Engineering and Advanced
Technology, 78-82.

1. The 45o orientation is the ideal reinforcement toresist flexural stresses followed by the 90O then
0O and lest for the URPB. The results showedthat the difference in means is significant for thedifferent types of reinforcement. Since the least
mean for the RPB which is 0O is significantly
higher than the mean of the URPB, it can beconcluded that the use of bamboo strips
reinforcement for the rice hull particle board
will really matter to increase the flexuralstrength of the particle board.

[2] Bystriakova, N., & Kapos, V. (2006). Bamboo
diversity: the need for a Red List review.
Tropical Conservancy , 12-16.
[3] Van der Lugt, P., van den Dobbelsteen, A. A.,& and Janssen, J. J. (2006). An environmental,
economic, and practical assessment of bamboo
as a building material for supporting structures.
Construction Building Materials .
[4] Rassiah, K., Ahmad, M., & & Ali, A. (2014).
Mechanical Properties of Laminated Bamboo
Strips from Gigantochloa Scortechinii/Polyester
Composites. Elseiver Materiald and Design , 1.

2. The Internal Bond Strength of the reinforced
particle boards were significantly different
having the 0O as the highest while the 45O is thelowest. Comparing the IBS mean of the RPB
with URPB, the latter is significantly higher
than the mean of the reinforced boards. Thisshows that if the quantity of bamboo strips
becomes greater, the result for IBS is lower.
Therefore the bamboo strips reinforcement does
not work for the internal bond strength.

[5] Pandey, C. N., & Sujhata, D. (2011). Crop
Residues, the Alternate Raw Materials ofTomorrow for the Preparation of Composite
Board. Indian Plywood Industries Research andTraining Institute.

[6]

3. The Impact Resistance mean of the reinforced
boards are higher than that of the URPB, though
there is no significant difference on the types ofreinforcement of the rice hull particle boards, aswell as the URPB. This shows that the use ofbamboo strips reinforcement does not
significantly affect the impact resistance of theparticle boards, though it may increase the IR
value.

Coxworth, B. (2015, July 10). Rice Husk may
Find Use in Cheaper, Greener, and Longerlasting
Particleboard:. http://www.gizmag.com/ricehusk-particleboard/37730/[7] Lagrimas, J. L., Mailom, M. C., & Matampale,
K. A. (2013). Development of Fiberboard with
Rice Hull as Base Material on Polystyrene as
Resin Binder. 20-28.

419

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STRENGTHENING EFFECT OF A CFRP ROD HAVING RIBSFOR A CANTILEVERED SLAB
Yusuke Kuroda1, Hiroaki Hasegawa2, Nobuhiro Hisabe2 and Isamu Yoshitake1
Department of Civil and Environmental Engineering, Yamaguchi University, Japan;
2
Mitsubishi Plastics Infratec Co., Ltd., Japan

1

ABSTRACTThe near surface mounted (NSM) method using carbon fiber reinforced polymer (CFRP) rod has been used tostrengthen bridge slabs. Bond between concrete and CFRP rod is a concern for strengthening concrete members
although the CFRP has superior structural performance. CFRP rods are manufactured by tensioning carbon fibers,
and coated with resin. Therefore the CFRP rod has relatively smooth surface and it has little mechanical shear
resistance. This study focuses on CFRP rods applied to cantilevered bridge deck slabs. The study examines
strengthening effect on reinforced concrete (RC) beams embedding various CFRP rods. To improve the bondperformance, CFRP rods coated with an epoxy adhesive incorporating silica sand (sand-coating), and CFRP rods
having artificial ribs made with glass fiber reinforced polymers (GFRP) sleeves were used in the experimental
investigation. Monotonic and cyclic loading tests were conducted to confirm the flexural behavior of the beamstrengthened with the NSM CFRP rods. In addition, finite element (FE) simulation was performed to quantify thebond effect of these CFRP rods.
Keywords: CFRP rod, Cantilevered Slab, Near Surface Mounted (NSM) Method, Bond, Fatigue

having artificial ribs made with glass fiber reinforced
polymers (GFRP) sleeves. Monotonic and cyclicloading tests were conducted to examine the flexural
behavior of cantilevered RC beam specimens
strengthened with the NSM CFRP rods. In addition,
finite element (FE) simulation was performed toquantify the bond effect of the CFRP rods. This paper
reports the experimental and numerical investigations.

INTRODUCTION
Cantilevered reinforced concrete (RC) slabs ofhighway bridges have been damaged by increasewind and traffic loads in recent year. Development ofstrengthening method for the upper surface of
cantilever slabs has been required. This paper
presents the near surface mounted (NSM) method
using carbon fiber reinforce polymer (CFRP) rod.
The CFRP rod has some advantages such as superior
fatigue durability and high chemical resistance.
Strengthening method for cantilever slab has beenstudied previous researches [1]-[7]. The NSM method
has already been employed in the world. CFRP
materials are categorized to two types which are highmodulus and high-strength. The CFRP rod used in the
study is high-modulus type because it has
significantly higher Young’s modulus than themodulus of rebar and can decrease stress ofreinforcing materials. The most concern of the CFRP
rods is low bond characteristic because the rod has no
mechanical shear connectors like rebars. Note is thatthe bond strength of CFRP rod is almost 40 % of thestrength of deformed bars. The purpose of this study
is to examine the strengthening effect for RC
members embedding CFRP rods. In particular, thestudy focuses on applicability of high modulus CFRP
rods for cantilevered slabs subjected to negative
bending moment. To improve bond properties, theexperimental investigation prepared two kind of
CFRP rods; a rod coated with an epoxy adhesive
incorporating silica sand (sand-coating); CFRP rods

EXPERIMENTAL PROCEDURE
Materials
Table 1 gives compressive strength and Young’s
modulus of concrete and high strength expansive
mortar at age of 28 days. Rebars having a nominal
diameter of 10 mm and high modulus CFRP rods ofmoderate tensile strength were used. Young's
modulus of HM12 is twice than rebar. The reinforcing
material contributes to reduce the stress of rebars.
Yield strength of the rebar is 345 MPa. Physicalproperties of the rebar and the CFRP rod are given in
Table 2.
Table 1 Properties of concrete and mortar

Compressive strength(MPa)
Young’s modulus (GPa)420

Concrete

Mortar

23.8

43.4

28.6

26.5

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 Properties of CFRP rod and rebar

Yielding strength (MPa)Tensile strength (MPa)Young’s modulus (GPa)

CFRP rod
N/A
1550
418

specimens of HM12-GS-300 were prepared and wereused for monotonic and cyclic loading tests,
respectively. The test specimens are summarized inTable 3.

Rebar
295
374
206

Table 3 Test specimens

Test specimens

Specimen I.D.
Control
HM12
HM12-EC

CFRP rod
Test
N/A
M
Type N
M
M
Type A
Type B
M, C
HM12-GS-300(interval of 300 mm)
Type B
HM12-GS-200
M(interval of 200 mm)
M: monotonic loading test; C: cyclic loading test

Figure 1 shows a schematic of test specimen. Fiveslab specimens were prepared to simulatecantilevered bridge deck slabs. The dimensions ofslab specimen were 200 mm deep x 250 mm wide x
1,850 mm long as shown in Fig. 1.

Flexural loading test

Fig. 1

Four-points loading test was conducted as shownin Fig.3. Span length and loading span were 1650 mm
and 250 mm, respectively. The load of 1.0 mm/min
was applied to the slab specimen in the monotonic test.
To examine fatigue durability, cyclic loading test
was also conducted by using the slab specimen of
HM12-GS-300. The maximum load in the fatigue test
was 60 % of the load-carrying capacity, and the cyclicload was applied with 1 Hz up to the 2 million cycles.

Detail of test specimen

Type NGFRP sleeve

CFRP rod (HM12)

Epoxy resin + silica sandRebar and CFRP rods were placed at 150 mm and
187 mm deep, respectively. All test specimens were
cured for 2 weeks using wet curing sheets. After that,the surface layer of concrete was removed by using awater-jet (water pressure: 220 MPa, water volume: 15
L/min). CFRP rods were placed on concrete whichremoved the surface layer. The concrete surface wasrefilled with high strength mortar. Schematics of
CFRP rods used in the test are shown in Fig. 2.

200mm
or
300mm

Fig. 3

Flexural loading test

RESULTS AND DISCUSSION
Type A

Type B

Monotonic loading test
Fig. 2

CFRP rods used in this test

Table 4 gives cracking load (Pcr), the maximum
load (Pmax) and failure mode of each specimen. The
maximum load of the specimen HM was 1.8 times
higher than the Pmax of the control specimen. In
addition, the maximum loads of HM12-EC, HM12GS-300, HM12-GS-200 were 2.7, 2.3, 2.8 times

Primary test parameters are reinforcing materials
such as CFRP rods (Type N), CFRP rods coated with
an epoxy adhesive incorporating silica sand (sandcoating: Type A), and CFRP rods having artificial
ribs made with GFRP sleeves (Type B). Two
421

GEOMATE- Bangkok, Nov. 14-16, 2016No.
higher than the Pmax of the control specimen,
respectively. The observation confirms that flexuralstrength of the strengthened beams is higher than thestrength of control specimen, high strengthening
effect of CFRP rods was confirmed in this test.
Test specimens strengthened with CFRP rods
were collapsed due to shear failure while the control
specimen indicated flexural failure. It is noteworthy
that the NSM of CFRP rods is sufficient forstrengthening of flexural member. It should be notedthat the NSM strengthening may induce shear fractureof CFRP rod rather than the flexural failure.
Figure 4 shows a typical shear failure. The failuremode of test specimen having GFRP sleeves (HM12GS-300 and HM12-GS-200) were almost similar, theshear crack generated from edge of GFRP sleeves. In
addition, interfacial failure of concrete-mortar layer
was observed in the test of strengthened specimens.
Figure 5 illustrates a typical failure of CFRP rod. The
failure was observed at the edge of GFRP sleeve. The
observation confirms that the CFRP rod has relatively
low resistance to shear while it has excellent tensile
strength.

1
2
3
4
5

Specimen I.D.

Pcr

Pmax

FailureControl
HM12
HM12-EC
HM12-GS-300
HM12-GS-200

(kN)
8.7
14.2
14.3
14.8
13.5

(kN)
38.0
68.6
101.4
87.0
105.6

mode
Flexural
F+S
Shear
Shear
Shear

Table 5 Maximum deflection and strain of CFRP rod
No.

Specimen I.D.

1
2
3
4
5

Control
HM12
HM12-EC
HM12-GS-300
HM12-GS-200

Deflection
(mm)
31.5
4.16
5.36
4.72
6.59

Strain of
CFRP rod
N/A
1165 x 10-6
1516 x 10-6
1554 x 10-6
1750 x 10-6

The maximum deflection and strain of CFRP rod
are summarized in Table 5. Also, Figure 6 shows
load-deflection responses. The result indicates thatthe maximum deflections of strengthened specimens
were significantly lower than the deflection of control
specimen. The observation confirms that the NSM of
CFRP rods increases ductility of the flexural member.
The deflection of strengthened specimens embedding
bond-improved CFRP rods was higher than thedeflection of HM12. The result confirms that CFRP
rods bonded firmly to concrete up to the shear failure.
120

Load （kN）

100

Fig. 4

Shear failure of test specimen

Control
HM12
HM12-EC
HM12-GS-300
HM12-GS-200

80
60
40
20
0
0

Fig.6

2
4
6
Midspan deflection （mm）

8

Load-deflection responses

Figure 7 shows load-midspan strain of CFRP rod.
Strains of CFRP rods having artificial ribs werehigher than the strain of HM12. The result also
confirms the improvement of bond performance of
CFRP rod having artificial ribs.
Fig. 5
Table 4

Failure of CFRP rod
Load bearing capacity422

GEOMATE- Bangkok, Nov. 14-16, 2016

120
1516 x 10-6

1554 x 10-680601165 x 10-6

60

40

HM12
HM12-EC
HM12-GS-300
HM12-GS-200

50

20

40
Load (kN)

Load (kN)

100

Figure 9 shows the load-midspan deflection response.
The deflection increased in accordance with loading
cycles.

1750 x 10-6

00Fig. 7

500
1000
1500
2000
Midspan strain of CFRP rod ( x 10-6)

30
20
10

Load-strain of CFRP rods00Cyclic loading test
A cyclic loading test using HM12-GS-300 was
carried out to examine fatigue durability of slab
specimen strengthened with bond-improved CFRP
rods. Four-points loading system was used as well as
the monotonic loading test. The maximum load was
designed as 60% of the load-bearing capacity of
HM12-GS-300. Herein, the minimum load was 10%of the static strength. At the number of 1, 10, 100,
1000, 10000, and 100000 cycle loadings, themaximum load was provided statically (3 kN/min.) to
the slab specimen.
Flexural cracks were firstly initiated near themidspan. The additional bending and shear cracks
were observed during the repeated loadings. The
shear crack was caused from edge of the GFRP
sleeves. It should be noted that the debonding
between concrete and mortar layer was not observedin the cyclic loading test. The typical fatigue cracks
are presented in Fig. 8.

Fig. 8

Fig.9

1

2
3
4
Midspan deflection (mm)5Load-midspan deflection response

Load / Deflection (kN/mm)

Figure 10 shows a transition of flexural rigidity
defined as load/deflection. Note is that the flexural
rigidity was remarkably decreased by 600,000
loadings and it was stable after the loading cycles of
600,000.
40
30
20
10
0
0

500
1,000
1,500
Fatigue cycle ( x 103)

Fig. 10

Transition of flexural rigidity

2,000

To examine residual strength of the slab specimen
subjected to the cyclic load, a monotonic loading test
was conducted after 2 million repeated loadings. The
slab was also collapsed by shear failure initiated fromthe fracture of CFRP rods. The residual loading test
indicated the load-bearing capacity strength of 110.4
kN. It is of interest that the residual strength was
higher than the static strength (87.0 kN) of the slab.
Figure 11 shows the crack patterns observed in themonotonic loading test and residual loading test.
Compared to the monotonic loading test, cracks in thecyclic loading test were distributed uniformly. The
crack distribution possibly contributed todeformation of the beam and reduced the stressconcentration of the CFRP rod.

Typical fatigue cracks

The repeated loading achieved 2 million cycles.

423

GEOMATE- Bangkok, Nov. 14-16, 2016of CFRP rods without GFRP sleeves was also used
this analysis as control to confirm the effect of sleeves.
Bond condition of control was supposed Type I.
Young’s moduli and Poisson’s ratio employed in the
FE simulation are given in Table 6.

Figure 12 shows load-deflection responses of
HM12-GS-300. The comparative result confirms thatthe reduction of flexural rigidity was negligible
because the slope of regression line in the residual testis almost equal to the slope in the static test.

a) Cracks in the monotonic loading test

Fig. 13 Analysis model of FE simulation
Table 6 Material properties for FE simulation: Cracks in the cyclic test
: Crack in the residual testb) Cracks in the residual loading test
Concrete
Mortar
HM-CFRP rod
Rebar
GFRP

Fig. 11 Crack pattern in the residual loading test120

yy = 18.779x - 25.472

Load (kN)100

y = 19.382x + 5.3436

Young’smodulus
(GPa)
28.0
29.5
500
200
30.0

Poisson’s
ratio
0.20
0.20
0.30
0.30
0.28

80

Table 7 Maximum strains (in FE analysis)

60
40

Residual test

20

Static test

00Fig. 12Type I
Type II

2

4
6
Midspan deflection (mm)8xHM12-GS-200
44.0 x 10-6
41.3 x 10-6

HM12-GS-300
43.9 x 10-6
39.9 x 10-6

Table 7 summarizes the maximum strainsobtained from the FE analysis. In addition, Figure 14
shows strain distributions of CFRP rod having GFRP
sleeves, control (without GFRP sleeves). The applied
load in the analysis is 10 kN in order to simulate
behavior of uncracked bridge deck slab. The result
reveals that the maximum strains obtained by FE
analysis are almost equal despite the different bondconditions. In addition, strain observed in themonotonic test (51.2 x 10-6 at 10.3 kN) was comparedto the theoretical strain in FE analysis. The
comparison implies that the linear FE modeling isappropriate to simulate the strain distribution. This
simulation result confirms the bond effect of themechanical shear connector of the GFRP sleeves(Type II) while strain distribution under unbond
condition is different compared to the full-bondcondition (Type I) and control.

Load-deflection responses

FE simulation
To examine the bond effect of CFRP rod withGFRP sleeves, finite element (FE) simulation was
performed. A beam model embedding CFRP rods
with GFRP sleeves (Fig.13) is used in the numericalinvestigation. The FE simulation uses different bondconditions for the GFRP sleeves (Types I and II).
Type I is a full-bond model which the CFRP rod and
GFRP sleeves interface is glued firmly to the mortar.
On the other hand, Type II is a model that includes
unbonded interface except for the blue surface (bond).
These models were used to compare the effect of
mechanical shear resistance of the GFRP sleeves
without chemical bond and/or friction. An FE model
424

GEOMATE- Bangkok, Nov. 14-16, 2016

0

Distance from the bar-end of CFRP (mm)250 500 750 1000 1250 1500 1750

0

The authors thank to Mr. Sakaki, a graduate
student of Yamaguchi University, for his support in
the numerical simulation.

Strain ( x 10-6)

10
20
30
40
50

Acknowledgements

51.2 x 10-6Type I
Type II
Control
Experiment

References[1] Firas A., Arnaud C., Raoul F. and Christian T.,

60“Strengthening of RC members with nearsurface mounted CFRP rods”, Composite
Structure, Vol.91, 2009, pp.138-147.

Fig. 14 Strain distribution of CFRP rod
CONCLUSIONSThis paper presents near surface mounted (NSM)
technique using carbon fiber reinforce polymer(CFRP) rods for cantilevered RC slabs. Monotonic
loading test and cyclic loading test were conducted byusing slab specimens strengthened with the CFRP
rods. FE analysis was performed to confirm bondperformance of CFRP rods having GFRP sleeves.
The conclusions of this study are as follows:(1) The load-bearing capacity of the slab specimen
strengthened with CFRP rods was higher than atleast 1.8 times to the capacity of the control
specimen. Improvement of bond performance of
CFRP rods having artificial ribs was confirmed.
Shear failures were observed in the test of
specimens strengthened CFRP rods.
(2) The fatigue test of slab specimen strengthened
with Type B CFRP rod (HM12-GS-300) was
achieved 2 million, and high fatigue durability
was indicated. Residual test showed the
maximum strength of specimen strengthened
with HM12-GS-300 (110.4kN) was higher than
the static load-bearing capacity of the slab
specimen under same condition.
(3) The maximum strains of CFRP rod in the FE
simulation were almost equal despite different
bond condition of the rod. The simulation
confirmed the effectiveness of the mechanicalshear connector of the GFRP sleeves (Type II).

[2] Lam L. and Teng J.G., “Strength of RC
cantilever slabs bonded with GFRP strips”,
Journal of Composites for Construction, Vol.5,
No.4, Nov. 2001, pp.221-227.
[3] Moataz B. and Khaled S., “Flexuralstrengthening of RC beams with prestressed
NSM CFRP rods - Experimental and analyticalinvestigation”, Construction and Building
Materials, Vol.23, 2009, pp.3292-3300.
[4] Oudah F. and El-Hacha R., “Fatigue behavior of
RC beams strengthened with prestressed NSM
CFRP rods”, Composite Structures, Vol.94,
2012, pp.1333-1342.
[5] Teng J.G., Cao S.Y. and Lam L., “Behaviour of
GFRP-strengthened RC cantilever slabs”,
Construction and Building Materials, Vol.15,
No.7, 2001, pp.339-349.
[6] Yoshitake I., Kim Y.J., Yumikura K. andHamada S., “Moving-wheel fatigue for bridge
decks strengthened with CFRP strips subject tonegative bending”, Journal of Composites forConstruction, Vol.14, No.6, 2010, pp.784-790.
[7] Yoshitake I., Kim Y.J., Yumikura K. and
Mimura Y., “Composite strips with various
anchor systems for retrofitting concrete beams”,
International Journal of Concrete Structures andMaterials, Vol.5, No.1, June 2011, pp.43-48.

425

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

CONSTITUTIVE MODELING OF COAL ASH USING MODIFIEDCAM CLAY MODEL
Erica Elice Saloma Uy1 and Jonathan Rivera Dungca1
1
Civil Engineering, De La Salle University, Philippines

ABSTRACT
Coal ash is a by-product from coal-fired power plants that generates electricity. These waste materials most of
the time is kept in storage facilities. Due to the growing demand in electricity generation, storage facilities canno longer accommodate the waste materials. Instead of disposing the waste materials, it can be used as aconstruction material. A promising civil engineering structure that can use coal ash as a construction material island reclamation. The structure needs huge volume of materials and this can maximize the usage of coal ash. In
the Philippines, most land reclamation projects are conducted at the sea. From this, distilled water was replacedwith sea water to have a better evaluation of the strength properties of coal ash. Consolidated drained test was
performed having three conditions with respect to sea water exposure. First condition is no exposure, second
condition is immediate exposure and third condition is prolonged exposure. Results show that coal ash exposed
to seawater, immediate and prolonged, has smaller shear strength. On the other hand, it still has reasonable
strength suitable for land reclamation projects. Constitutive modeling using Modified Cam Clay model isincorporated in the study to be able to predict its behavior and failure in terms of mean effective stress, deviator
stress and specific volume.
Keywords: Coal Ash, Critical-State, Modified Cam Clayrequirements for highway embankments [5].
Another study showed its applicability forlightweight concrete production [6]. A promising
civil engineering structure that can use the byproduct as construction material is land reclamation.
This civil engineering structure is normallyconstructed in areas where there is a shortage inspace. This type of structure is usually constructed in
the Philippines especially in highly urbanized areas.
It is the objective of this study to determine theperformance of the by-product as a construction
material. To be specific, bottom ash was tested
under Consolidated Drained Triaxial Test. The effectof seawater is considered so distilled water waschanged to seawater. Three conditions were
simulated in the test, first condition is no exposure,second condition is immediate exposure and third
condition is prolonged exposure. Modified Cam
Clay model is incorporated to study its behavior
with respect to the mean effective stress (p’),
deviator stress (q) and specific volume (v).

INTRODUCTION
The Philippine government relies on coal-fired
power plant to provide electricity. The country
already has 19 power plants, 11 in Luzon, 6 in
Visayas and 2 in Mindanao [1]. The Department ofEnergy recently announced that 23 new power plants
will be opened by 2020 [2]. This is to address theincreasing demand of electricity of the country.
These power plants produce by-products such as fly
ash and bottom ash. They are normally stored at
impoundments or landfills. There are cases that thecapacity of the storage facility is not enough and thisleads to problems in proper disposal of the wastematerials. This can have a negative impact to theenvironment and community close to the storage
facility. In the Philippines, this incident is possible
since coal-fired power plant is one of the main
sources of electricity. It provides approximately 40%of the total power generation needed by the country[2]. Furthermore, most power plants operate for 24hours such as the coal-fired power plant in Calaca,
Batangas. The power plant is estimated to discharge
at about 62.62 tons of ash per hour [4]. Instead ofdisposing or storing the by-products it is better tofind an alternative use such as a construction
material. There are already existing studies thatthese by-products have a potential in substituting
some construction materials for civil engineering
structures. A study on its applicability as a sub basematerial for highway embankments was conducted.
Based on the results, it meets the strength

BOTTOM ASH
The bottom ash samples were obtained from
IloIlo Power Plant. From visual inspection, thematerial had three compositions. Particles with thesize of small gravels, sand, and fines were present.
For the fines, it is lightweight which makes it easily
blown away by air. Its index properties are
determined similar to [3] and the results are426

GEOMATE- Bangkok, Nov. 14-16, 2016site conditions [8], [9]. The target value is 95% since
it was the desired in situ condition. The maximum
dry density served as the reference point to achieve
the target relative compaction. The amount ofdistilled water or sea water was based on the valueof optimum moisture content.
Three conditions were prepared to check theeffect of seawater towards the strength of bottom ash.
The first one is the sample with no exposure. For
this case, only distilled water was mixed with thebottom ash. The second condition is the immediate
exposure to sea water. For this case the sample wasstill prepared similar to the first condition. It isduring the Consolidated Drained Test where thesample was exposed to sea water. For the third
condition, prolonged exposure was simulated so sea
water was already mixed with the bottom ash prior
to the Consolidated Drained Test. The amount of sea
water is computed based on the optimum moisture
content. The samples for the three conditions were
soaked for 16 hours as per ASTM provisions.

tabulated in Table 1. For the index properties,
bottom ash is classified as silty sand (SM). It has noplasticity since liquid limit and plastic limit is zero.
It is noticeable that the sample has sufficient amountof sand particles and followed by fines content.
Compared to other studies, the result of the specificgravity is higher than the range of values suggested,
1.899 to 1.903 [7]. Their coal ash was from adifferent manufacturing plant. This might havecontributed to the difference in index properties of
the coal ashes.
Table 1. Index properties
Index Property
Specific gravity(Gs)Liquid limit(LL)
Plastic limt (PL)Maximum void ratio (e max )Minimum void ratio (e min )Maximum dry unit weight(γ dmax )Optimum water content (ω opt )
USCS

2.25
0.00%
0.00%
0.94
0.85
13.94 kN/m3
15.85%SM%Gravel

0.86

%Sand

50.44

%Fines

48.70

Sea water Preparation
Artificial sea water was used in the experiment.
It was formulated following ASTM D 1141 – 98,
also known as the “Standard Practice for the
Preparation of Substitute Ocean Water”. A pH valueof 8.2 and chlorinity of the sea water was maintained
at 19.38. The chemical composition of the sea wateris shown in Table 2.

Table 2. Chemical composition of sea water
CompoundConcentration,g/L
NaCl
24.53
MgCl
5.2
NaSO
4.09
CaCl
1.16
KCl
0.695
NaHCO
0.201
KBr
0.101
HBO
0.027
SrCl
0.025
NaF
0.003
Ba(NO
9.94E-05
Mn(NO3.40E-05
Cu(NO3.08E-05
Zn(NO9.60E-06
Pb(NO6.60E-06
AgNO
4.90E-07

Consolidated Drained Test
Consolidated Drained Test was performed toevaluate the strength of the sample against long-termloading. The test has three phases namely, saturation,
consolidation and shearing phase. The saturation
phase was first performed to ensure that the sample
is fully saturated. The B-value or the ratio of thepore water pressure and confining pressure wasconstantly monitored. This parameter must have avalue greater than 0.95 to ensure a fully saturated
sample. The confining pressures (σ 3 ) used toconsolidate the sample are 50 kPa, 100 kPa, and 200
kPa. They are applied incrementally to have aconservative simulation of the stresses. The sample
was allowed to consolidate until there are no
significant volume changes in the sample. For the
shearing phase, a slow rate of 0.05 mm/min wasused to ensure the dissipation of pore water pressure
is attained. This phase usually lasts for a minimumof 4 hours to a maximum of 7 hours. For the first
condition, 3 samples were tested for each confiningpressure. For the second condition, 2 samples were
tested for each confining pressure. Lastly for thethird condition, 1 sample was tested for eachconfining pressure. From this, a total of 18 samples
were tested for this experiment.

EXPERIMENTAL PROGRAM
Triaxial Compression Phase
Sample Preparation
The sample was prepared by moist tamping.
Relative compaction was the controlled parameter toachieve a target initial condition. This was usedinstead of relative density in order to simulate actual427

GEOMATE- Bangkok, Nov. 14-16, 2016

RESULTS AND DISCUSSION
Stress-Strain Behavior and Mode of Failure

Deviator Stress, q (kPa)

The stress-strain plots, Fig. 1, presented in thissection were based on the sample’s critical state or at
failure condition. Based on the results, it wasobserved that all trials for each condition had asimilar trend. Therefore, the results presented in thissection are based on the last trial.
When no exposure to sea water was tested, it
can be seen that at 200 kPa the sample experienced
yield at less than 0.1% axial strain. At this point, thesample already exceeded its elastic limit. Following
this behavior, the sample reached its plastic state as
it is being axially compressed. Strain hardening ismore pronounced after the yielding of the soil. Thisbehavior is only visible in this condition. A common
trend from all the condition is no defined peak canbe seen in the stress-strain plot. Furthermore, thedeviator stress and axial strain increases as theconfining pressure was increased. The no exposure
to sea water condition and prolonged exposure to sea
water had almost the same results when it comes tothe values the deviator stress experienced. The axial
strain for no exposure to sea water had a larger value.
For the immediate exposure condition, the deviator
stresses are smaller. Based on the stress-strain
behavior, prolonged exposure to sea water has
minimal impact to the behavior of the bottom ash
since it arrived at a similar trend with the first
condition.
For its mode of failure, shear failure wasobserved for all tests as seen in Fig. 3. It is expected
when the sample is a brittle sample. Only samples
that are sandy can experience this type of failure.
From this, the bottom ash used can be classified as abrittle sample. Furthermore, the presence ofsufficient amount of sand in its composition
contributed to this type of failure.

0

0.4

Deviator Stress, q (kPa)

500
400
300
50kPa
100kPa200kPa200
100
0Deviator Stress, q (kPa)

0
700
600
500
400
300
200
100
0

0.1
0.2
0.3
Axial Strain, mm/mm
(b)
50kPa
100kPa200kPa

0.4

0

0.1
0.2
0.3
0.4
Axial Strain, mm/mm(c)Fig. 1Stress-strain plot of Bottom Ash a) no
exposure b) immediate exposure and c) prolonged
exposure.

The critical state parameters were establishedbased on the critical state model. The slope, M, ofthe critical state line (CSL) with respect to thedeviator stress and mean normal effective stress was
first established. This parameter can be obtained
from Fig. 3. The area below CSL is the safe zone or
the stresses where the sample is still in its elastic
state. Based on the results, the second condition had
the smallest slope, 1.3531. The third condition has
the largest value of 1.4914. The slope has a relation
to the effective critical angle of internal friction(φ’ crit ) as shown [10]:

3M
6+M

0.1
0.2
0.3
Axial Strain, mm/mm
(a)

600

Critical-State Parameterssin ϕ 'crit =50kPa
100kPa200kPa

700
600
500
400
300
200
100
0

The resulting values are as follows, 33.51°,
35.26°and 36.67° for the first, second and third
condition, respectively. These values are the typicalresult of silty sand [11].
The equations for isotropic normal compression
line (NCL) are also established to determine thecritical state line of the bottom ash, Table 3. The
equations resemble the equation of the line. From
this, the critical state parameters such as slope ofNCL (λ), y-intercept of NCL (N p ) and y-intercept of
CSL (Γ) can also be established from the equations
and are presented in Table 4. These parameters areused to plot create CSL with respect to the deviator
stress and the mean normal effective stress. A
sample plot is shown in Fig. 4. The points of NCL

(1)
428

GEOMATE- Bangkok, Nov. 14-16, 2016

are from the consolidation phase while the points
along CSL are from the shearing phase. The area
beyond NCL is the impossible state. The area below
NCL is where over-consolidation is encountered. On
the other hand, the area below CSL is where
expansion happens.

Deviator Stress, q (kPa)Mode of failure for bottom ash.

The NCL and CSL for condition 1 and 2 werecombined as seen in Fig. 5 and 6. Different
behaviors can be observed from the 2 conditions. It
is expected that as the confining pressure isincreasing the specific volume is also increasing. For
the first condition, the specific volume for 50 kPa
and 200 kPa are almost similar. On the other hand,for 100 kPa, its specific volume decreased
dramatically. The particle rearrangement for thisconfining pressure is larger compared to 50 kPa and200 kPa. For the CSL, same observations can beseen for the specific volume. For the mean normal
effective stress, 200 kPa had the largest values. For
the second condition as seen in Fig. 6, the values of
specific volume of NCL for 50 kPa and 100 kPa arealmost the same. It can be observed that 50 kPa hadslightly higher values. This is also evident for the
CSL plot. The exposure to sea water may haveaffected the results of NCL and CSL of the second
condition. The sample were 100 kPa was applied
seemed weaker than the other samples for second
condition.
Isotropic

normalcompression

Immidiate Exposure

50kPa

2.2149 - 0.0954lnp'

1.9657 - 0.0602 lnp'

100kPa2.8058- 0.2736 lnp'

1.9548 - 0.0684 lnp'

200kPa2.1871- 0.0845 lnp'

2.4674 - 0.0922lnp'

Np

Γ

50kPa
100kPa200kPa50kPa
100kPa200kPa2.21492.80582.1871
1.96571.95482.46742.11952.53222.10261.90551.88642.3752

Immediate
No
Exposure Exposure

Table 4. Critical state parametersσ3λ0.09540.27360.08450.06020.06840.0922

600

CSL
50kPa
100kPa200kPa

500
400
300

Deviator Stress, q (kPa)q = 1.3531p'
R² = 0.9210

200
100
0

700
600
500
400
300
200
100
0

100
200
300
400
Mean Normal Effective Stress, p' (kPa)(b)
q = 1.4914p'
R² = 0.9893

CSL
50kPa
100kPa200kPa
0

200
400
600
Mean Normal Effective Stress, p' (kPa)(c)
Fig. 3 Critical state line with respect to thedeviator stress and mean normal effective stress a)
no exposure b) immediate exposure and c)
prolonged exposure.

line

No Exposure

100 200 300 400 500
Mean Normal Effective Stress, p'…
(a)0Specific Volume, v

Table 3.
equations
σ30Deviator Stress, q (kPa)

Fig. 2CSL
50kPa q = 1.4299p'
100kPa R² = 0.9923
200kPa

700
600
500
400
300
200
100
0

1.9
1.85
1.8
1.75
1.7
1.65
1.6
1.55

NCL
CSL

0

500
1000
1500
Mean Normal Effective Stress, p' (kPa)

Fig. 4 Critical state line with respect to thedeviator stress and mean normal effective stress a)
no exposure b) immediate exposure and c)
prolonged exposure.
429

kPa. Only the equations with 200 kPa are presented
since it can cover all the stresses from 50 kPa and100 kPa. It can be seen that Eq. 5 had the smallest
preconsolidation stress and slope M.

2
1.8
1.6
1.4
1.2
1
0.8
0.6
0.4
0.2
0

2.5
50kPa
100kPa200kPa
0

500
1000
1500
Mean Normal Effective Stress, p' (kPa)(a)

Specific Volume, v

Specific Volume, v

GEOMATE- Bangkok, Nov. 14-16, 2016

2
1.5
50kPa

1

100kPa0.5

200kPa

0
0

500
1000
1500
Mean Normal Effective Stress, p' (kPa)

1.5

(a)
2

1

50kPa

0.5

100kPaSpecific Volume, v

Specific Volume, v

2

200kPa
0
0

200
400
600
Mean Normal Effective Stress, p' (kPa)(b)
Fig. 5 Specific volume versus mean normal
effective stress plot for no exposure to sea water a)
NCL and b) CSL.

1

50kPa
100kPa0.5

200kPa
0
0

100
200
300
400
Mean Normal Effective Stress, p' (kPa)(b)
Fig. 6 Specific volume versus mean normal
effective stress plot for immediate exposure to sea
water a) NCL and b) CSL.

Modified Cam Clay Model
Modified cam clay model (MCC) is an elastoplastic model that characterizes the soil in terms ofthe mean normal effective stress, deviator stress andspecific volume [12]. The model is the result of theenhancements made in the Cam Clay Model. The
model has an elliptical yield locus having thisequation [10]:p'=(1013.16p'=(
960.272

p'M=
p 'c M 2 + q

1.5

(3)

p'
where:
p’ = mean normal effective stressq = deviator stressp’ c = preconsolidation stress
M = slope of CSL from q vs p’

1.429922
1.4299 + (q

)

2

p '2

1.353121.35312 + (qp'=(1030.02

)
)

2

(5)

)
p '2

1.491421.49142 + (q

(4)

)

2

)
p '2

(6)

The yield locus presented in Fig. 7 serves as theboundary of the soil with respect to plastic andelastic state. A similar trend was observed for all
samples for the plot of yield locus. Points within theyield locus are stresses at the elastic state while
points beyond are at its plastic state or failure state.

The preconsolidation stress is estimated byusing Eq. 3. The q and p’ at critical state was used[10]. The largest preconsolidation stress bottom ash
can withstand is 1030.02 kPa. This value wasobtained from the third condition. The equations
presented were used to create the yield locus seen in
Fig. 7 for a sample with a confining pressure of 200
430

GEOMATE- Bangkok, Nov. 14-16, 2016

Deviator Stress, q (kPa)

CSL

ESP

Initial YL

MCC[2] Gonzales, I. C. (2015). 23 new power plants to
go online by 2020. The Philippine Star.
Retrievedfrom
http://www.philstar.com/business/2015/06/15/1
465896/23-new-power-plants-go-online-2020.
[3] Dungca, J. R. & Dychangco L. F. T. (2016).
Strength properties of road base materials
blended with waste limestones. International

1400
1200
1000
800
600
400
200
0
0

500
1000
1500
Mean Normal Effective Stress, p' (kPa)

Fig. 7
Stress path with yield locus of bottom
ash CD test with 200 kPa confining pressure.
CONCLUSIONS
Bottom ash was tested under Consolidated
Drained Triaxial Test. The objective of the study
was to investigate on the possibility of using the byproduct as a material for reclamation. Exposure tosea water was considered since most reclamation
projects are situated at the sea. Three conditions
were considered, no exposure, immediate exposure
and prolonged exposure. Results show that
prolonged exposure to sea water has a minimal
difference when compared to the no exposure
condition. The values of effective critical angle ofinternal friction are as follows 33.51°, 35.26°and
36.67°. The value is comparable to a silty soil were
its typical values ranges from 26° to 35°. For the
Modified Cam Clay Model, the preconsolidationstress was computed and a similar trend wasobserved. The computed maximum value is 1030.02
kPa. The value is from the prolonged exposure
condition. From this, bottom ash is a promisingmaterial that can be used in a reclamation project.
REFERENCES[1] Department of Energy. (2016). List of existing
power plants as of December 2015. Retrieved
from
http://www.doe.gov.ph/power-andelectrification/list-of-existing-power-plants.

431

Journal of Geomate. 11(25). 2493-2498[4] Antonio, O. (2006). Lime-fly ash stabilizationof Calaca, Batangas bottom ash. [Graduate
Thesis] University of the Philippines, Quezon
City, Philippines.
[5] Jao, J. (2012). Strength and permeability
characteristics of road base materials blended
with fly ash and bottom ash. [Graduate Thesis].
De La Salle University, Manila, Philippines[6] Ahmaruzzaman, M. (2010). A review on theutilization of fly ash. Progress in Energy andCombustion Science, 36(3), 327-363. Elsevier
Ltd. doi:10.1016/j.pecs.2009.11.003[7] Antonio, O. V. M., & Zarco, M. A. H. (2007).
Engineering properties of Calaca Batangas
bottom ash, 28(1), 37–56.
[8] Zabielska-Adamska, K. (2006). Shear strengthparameters of compacted fly ash–HDPE
geomembraneinterfaces.
Geotextiles
andnGeomembranes,n24(2),n91–102.
doi:10.1016/j.geotexmem.2005.11.006.
[9] Kim, B., Prezzi, M., & Salgado, R. (2005).
Geotechnical properties of fly and bottom ash
mixtures for use in highway embankments, 914924.
[10] Powrie, W. (2004). Soil Mechanics Concepts
and Applications Second Edition. London:
Spoon Press.
[11] Das, B. & Sobhan, K. (2014). Principles ofGeotechnical Engineering Sixth Edition, UnitedStates: Chris Carson.
[12] Munda, J., Pradhan, P. K. & Nayak, A. K.
(2014). A Review on the performance ofModified Cam Clay Model for fine grained soil.
Journal of Civil Engineering and Environmental
Technology. 1(5), 65-71.

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ASSESSMENT OF CRITICAL-STATE SHEAR STRENGTH
PROPERTIES OF COPPER TAILINGS
Erica Elice Saloma Uy1 and Mary Ann Q. Adajar1
Civil Engineering, De La Salle University, Philippines

1

ABSTRACT
Philippines, with a total of 7,107 islands, have one of the largest mineral resources in the world. The copperand gold deposits are considered to be among the largest in the world (Bureau of East Asian and Pacific Affairs,2010). Mining minerals generate waste material called mine tailings. Impoundment of mine tailings is normallydone to store these waste materials. One of the most common types used in impounding tailings is raised
embankments because of its low economic cost. This impoundment uses natural soil, tailings, and waste rocks as
the construction material. There are cases that raise embankments experience stability failure which can affectthe environment and well-being of the community where it is situated.
It is the interest of this study to assess the possibility of using mine tailings, specifically copper tailings, as aconstruction material. Index properties were first established following ASTM standards. From this, it wasestablished that the copper tailings has plasticity. Unconsolidated undrained, consolidated undrained andconsolidated drained test were conducted to determine the critical shear strength of the copper tailing. The
copper tailings were tested having two relative densities namely, 60% and 90%. The effective critical angle offriction was found to have a range of 21°-28°. Since critical-state parameters are considered in the study, Camclay model can be implemented so that to its behavior and failure mechanism can be predicted.
Keywords: Mine Tailings, Critical-State, Cam Clayenvironmentally responsible mining. Up to this date,there have been 21 tailing dam failures in thePhilippines from 1982 up to 2007. Six out of 21
events were classified as major dam failure [3].
These events harmed the communities living near
the mining facility. It also polluted the bodies ofwater destroying the livelihood of the locals. Sincemine tailings are valueless and are produced in large
volumes, it is normally stored at impoundment ofmine tailings. Instead of just storing the wastematerials, this study aims to investigate its potential
as a construction material. Even though there are
already several studies performed in other countries,
there is a lack of research in the Philippines
regarding the geotechnical characteristics of minetailings [2].
It is the objective of this study to establish theindex properties and critical shear strength of coppermine tailings. Index properties were establishedfollowing American Society for Testing andMaterials (ASTM) standards. Critical state condition
is considered to have a better understanding of themechanical behavior of the copper mine tailings.
Unconsolidated undrained (UU), consolidated
undrained (CU) and consolidated drained (CD) test
were performed using a triaxial apparatus todetermine its critical shear strength. Cam clay modelis incorporated in the study since critical state
condition is applied.

INTRODUCTION
Philippines, with a total of 7,107 islands, haveone of the largest mineral resources in the world.
According to National Economic Development
Authority (NEDA), the country has an estimated
USD 840 billion (Php 47 trillion) worth of
unexplored gold, copper, nickel, chromite,
manganese, silver and iron [1]. The copper and gold
deposits of the Philippines are also among thelargest in the world. Mining generates wastematerial called mine tailings. The waste materials
are commonly stored by impoundment of the minetailings. A raised embankment is widely adopted in
impounding tailings due to its low economic cost [2].
This storage facility uses natural soil, tailings, andwaste rocks as the construction material. There arecases that raise embankments experience stability
failure which can affect the environment and wellbeing of the community where it is situated. One of
the well-known disasters in the Philippine mining
industry was the Marcopper incident whichhappened in 1996 [3]. Mine tailing spill occurredbecause of the leak at the drainage of the storage
facility the Tapian Pit. This caused a fracture in thedrainage tunnel which resulted to the spill thatconsists of four million metric tons of waste.
Although Philippines is gifted with vast amount ofminerals, not all companies follow the framework of
responsible mining. One of which is in accordancewith the international best practice such as ensuring432

GEOMATE- Bangkok, Nov. 14-16, 2016

loose, medium dense and highly dense reconstituted
sample, respectively. The amount of water mixed
with the sample was less than its optimum moisture
content. This allows the sample to be molded
without crumbling. The sample was soaked for 16
hours as per ASTM standards. A cylindrical moldwith a diameter of 38 mm and a height of 76 mm
was used. This mold was used to ensure that thediameter of the sample was at least 30 mm. The
largest particle which was contained within thesample was also maintained to be 1/6 smaller of its
diameter. A ratio of the height to diameter between 2
to 2.5 is followed [6].

COPPER MINE TAILINGS
Copper mine tailings are obtained from Barrio
Maglinao, Municipality of Basay Negros Oriental,
Central Visayas. The mining site was operated byConstruction and Development Corporation in thePhilippines (CDCP) in 1979 and it was closed downin 1983 [4]. The site left negative environmental
impacts to the Municipality of Basay especially to the
Pagatban River. The mineralization for the Basay
Copper Project is copper porphyry. Extensive open
mining and limited underground mining was
conducted between 1979 and 1983. In order toextract copper, the ore was first crushed and grinded.
Forth floatation was performed to remove unwanted
rock, gangue, from the copper ore. From this process,a 25% copper concentrate was achieved. Further
refinement of the mineral was performed by smelting.
The impurities were removed in a form of a slag. The
mineral would now be in copper iron sulphidesolution or matte. This is followed by conversion
from matte to blister copper or partly purified copper
having a blistered surface. In preparation for the laststage of extraction, the blister is casted into anodes.
Lastly, electrolytic refining is executed by purifying
the copper up to 99.99% through electrolysis [4]. The
waste generated from extraction, copper mine tailings,
were stored in tailing dams. The copper tailings
obtained were in a somewhat dry condition and it
formed a rock like formation.

Unconsolidated Undrained Test
The unconsolidated undrained (UU) test or Quick
Test (Q-Test) was performed based on ASTM D
2850. For this set-up, drainage was not permitted all
throughout the experiment. Consolidation was notnecessary but confining pressures are still applied ata fast rate. The rate of loading used for applying theaxial load is dependent on the type of soil tested. For
this study, a rate of approximately 1%/min is usedsince the soil is has plasticity. It was stopped when
the sample experiences failure or it reached criticalstate. For the experiment proper, samples wereprepared having an initial relative density of 27%,
60%, and 90%. The confining pressures (σ 3 ) usedare 25 kPa, 50 kPa and 100 kPa [6]. A total of 9
samples were tested for the UU-condition.
Consolidated Undrained Test
The consolidated undrained (CU) test or R-Test(Rapid Test) was performed based on ASTM D
4767 – 02. For this condition, the specimen wasallowed to consolidate prior to the shearing phase.
There are three phases in this condition namely,
saturation, consolidation and shearing phase. For thesaturation phase, to ensure that the sample is fully
saturated a B-value is computed. It is the ratio of thepore water pressure and confining pressure at thetime when the volume change is constant. The Bvalue must be greater than 0.95 to ensure that the
sample is fully saturated. For the consolidation
phase, the confining pressure (σ 3 ) was applied. The
drainage was left open until zero pore water pressureis achieved [8]. For the shearing phase, the rate ofloading of 1%/min was applied to simulate rapid
loading. It was stopped when failure or critical state
condition was experience. For the experiment proper,
samples were prepared having an initial relative
density 60%, and 90%. The confining pressures usedin this condition are 50 kPa, 100 kPa and 200 kPa[9]. The 25 kPa confining pressure was not appliedin this condition because the range of back pressuresused is larger. The sample will not consolidated
instead it will saturate. A total of 6 samples were
tested for the CU-condition.

EXPERIMENTAL PROGRAM
Preliminary Phase
The index properties of the copper mine tailings
were obtained in accordance to ASTM standards.
The following are the tests performed: MechanicalSieve Analysis (ASTM D422), Hydrometer Test
(ASTM D422), Atterberg limits [Plastic, Liquid(ASTM D4813) and Shrinkage limit (ASTM D427)],
Specific Gravity (ASTM D854), Compaction Test
(ASTM D698), Relative Density Test and Unit
Weight Determination which includes the Maximum
Index Density (ASTM D4253) test and Minimum
Index Density Test (ASTM D4254).
Triaxial Compression Phase
Sample Preparation
The reconstituted samples were prepared bymoist tamping. The method was applied since it
minimizes particle segregation and it is applicablefor most types of sands in compacting under a wide
range of relative densities. The samples were
initially dried to accurately achieve the desired
relative density (D r ) of 27%, 60%, and 90% [5]. The
mentioned relative densities are used to simulate a433

GEOMATE- Bangkok, Nov. 14-16, 2016

Consolidated Drained Test
The consolidated drained (CD) triaxial test or STest (Slow Test) was based on the British Standard
Methods of test for Soils for Civil Engineering
Purposes. The test’s purpose was to establish thedrained shear strength parameters such as the criticalstate effective angle of internal friction (φ’ crit ). For
this condition, water was allowed to drain during theshearing phase. This was to simulate and determine
the strength of the soil with respect to long-termloading. Similar to the CU test, CD test has threephases. The saturation and consolidation phase aresimilar with CU test. The rate of loading for the
shearing phase must be slow enough to allow thedissipation of excess pore water pressure. A rate ofloading of 0.5% axial strain per minute was usedsince it is applicable for sandy soil [15]. The
confining pressure used for the consolidation phase
are 50 kPa, 100 kPa and 200 kPa were used.

performed in order to create the GSCD. The sample
was sieved between #4 and #200 sieve. It resulted to
the following composition, 0.15% gravel, 59.04%
sand and 40.81% fines.
To properly define the soil classification of thecopper tailings, the Unified Soil Classification
System (USCS) was used since it can be applied in
all types of soil [10]. Also, the system can further
classify the soils into coarse-grained and finegrained soils. From the system, the copper tailings
can be classified as silty sand. Another classification
system was used to check the suitability of copper
tailings as support for roadways. The American
Association of State Highway and Transportation
Officials (AASHTO) was used since it can classify
the material’s general subgrade rating for roads.
According to AASHTO soil classification system,
the soil sample is classified as A-4 with a groupindex of 0. A group index close to zero indicates it isa good soil for highway subgrade. Its generalsubgrade rating is fair to poor. For this rating, thecopper tailings can have fair to poor drainage andsupport when used as a subgrade material forembankments [10]. The results of the index
properties were compared to some existing
researches of copper tailings. The result of thespecific gravity was compared to the research
Characteristics of Copper Mine Tailings: A Case
Study. Their sample was from the Sarcheshmeh
Copper Complex Tailings Dam in Iran. Its specificgravity ranges from 2.6 to 2.8 [11]. The copper
tailings tested have a larger value, 2.82, and its
percentage error is 0.71%. The result of theshrinkage limit was also compared to the researchLaboratory Properties of Mine Tailings. Theirsample was from Kennecott Mining in Salt Lake
County, Utah. Its shrinkage limit is 24.4% [9]. The
shrinkage limit of the copper tailings is smaller,
21.58%, and its percentage error is 0.12%. The soil
classification of the copper tailing from Kennecott
Mining is also silty sand but the fines content isgreater. The samples from Kennecott Mining only
have 31.3%. Results from other researches may varysince the samples are taken from different areas. The
origin of the sample may have an effect to the overall composition of the sample.

RESULTS AND DISCUSSION
Index Properties
The index properties of the copper tailings arepresented in Table 1. The liquid limit falls under the
category of a silty soil since it is in between 30-40%.
The plastic limit, when computed, is characterized asa clayey soil based on the range of 25-50% [5]. The
specific gravity is within 2.6 to 2.9 which is therange of values for a clayey and silty soil [7]. A
grain size distribution curve (GSDC) wasestablished as seen in Fig. 1.
FINES

SAND

GRAVEL

Percentage Passing, %

100
80
60
40
20
0
0.001

Fig. 1

0.01

0.1
Diameter, mm

1

10Grain size distribution curve of copper
tailings.

Stress-Strain Behavior and Mode of Failure
The copper tailings were sheared until failure oronce critical state is achieved. For the three triaxial
tests a common trend was observed from the stressstrain plot. A sample result is presented in Fig. 2.
Yield was constantly experienced during the earlystages of the shearing phase. Based on the plots it
normally happens at less than 5% strain. Thisbehavior shows that the sample had alreadyexceeded its elastic limit and immediately shifted toa purely plastic sample while it is being sheared. It is

The plot is used to determine the soil gradationof the copper tailings. It contained a flat section at
the fines section which is the indication that the
sample is gap graded. A gap graded soil usuallycontains a large percentage of big and small particles
with a limited amount of intermediate sizes [12].
When this type of gradation is compacted, it can
result to larger voids compared to a well graded soil.
The composition of the copper tailings was also
determined through the sieving process that was434

GEOMATE- Bangkok, Nov. 14-16, 2016that Su ave at 90% relative density decreased. This
result is due to the sufficient amount of fines present
and a quick test was performed. Copper tailings arecomposed of 40.81% fines which made it a rather
weak sample at this relative density and test [8].
The critical undrained shear strength (Su crit ) was
obtained from the CU test. Results are tabulated atTable 3. It can be observed that a different trend
occurred in this set-up. The samples prepared with adenser composition, D90, had a larger range ofvalues. The void ratio after the consolidation wascomputed as seen in Table 4. The void ratio isrelatively smaller after consolidation whencompared to the initial void ratio. A smaller void
ratio translates to a higher relative density. This
explains the trend experienced in for the Su crit .
The consistency of the sample can be classified
based on its undrained shear strength [8]. The
Copper tailings’ Su ave is classified as a sample withmedium consistency. For the Su crit it is classified asa sample with medium to very stiff consistency. Achange in consistency can be observed because theconsolidation phase of CU test improved theresistance of the sample to compression. A medium
to very stiff consistency would have an undrained
shear strength ranging from 25 kPa to 200 kPa [8].

followed by strain softening for a small amount oftime then it shifts to strain hardening all throughout.
No peak was observed in the plots which can result
to a plastic mode of failure. This was verified during
the experiment that the sample had experienced abulging mode of failure in all tests, Fig. 3. Based on
the observed stress-strain behavior and the mode of
failure, this indicated that the sample tested is ductile.
Table 1. Index properties
Index Property
Specific gravity(Gs)Liquid limit(LL)
Plastic limt (PL)
Plasticity index(PI)Shrinkage limit(SL)Shrinkage Ratio(R)Maximum void ratio (e max )Minimum void ratio (e min )

Value
2.82
32.05%
25.81%
6.24%
21.58%
1.65
1.441
0.964

Maximum dry unit weight(γ dmax )15.70 kN/m3

Optimum water content (ω opt )19.47%

Deviator Stress , q (kPa)

140
120
100
80
60

d60 25kPa
d60 50kPa
d60 100kPa

40
20

Table 2. Average undrained shear strength (Su ave )Dr
CopperTailings

00

Fig. 20.1
0.2
0.3
Strain, mm/mm

0.4

Stress-strain plot of copper tailings for UU
test Dr = 60%

D27
45.39

Su ave (kPa)
D60
46.04

D90
32.65

Table 3. Critical undrained shear strength (Su crit )Su crit (kPa)Dr
Copper Tailings

D60
42.42-74.12

D90
78.49-117.75

Critical-Shear Strength

Fig. 3The critical shear strength is measured bydetermining the critical angle of internal friction
from the CU and CD test. The typical values forangle of internal friction for silts range from 26° to
35° [8]. Based on the results as seen in Table 5, only
the values at 60% relative density for both CU and
CD test are in close agreement. The results are also
compared to the shear strength of Copper Tailings
from Kennecott Mining which is 34° [9]. Copper
tailing’s strength is considerably smaller since it has
a greater amount of fines content, 40.81%. The
samples from Kennecott Mining only have 31.3%.
Based on the results of shear strength parameters,
Copper tailings has sufficient strength but it is not
close to the typical materials used for subgrade,embankment or structural fill. The typical materials
used are well-graded sands with greater than 34°
angle of internal friction [16].

Stress-strain plot of Copper Tailings

Undrained Shear Strength
Average undrained shear strength (Su ave ) was
computed from the three confining pressures appliedin the UU test. The results are tabulated at Table 2.
The average value was obtained since the deviator
stress from the Mohr’s Circle was not equal. Resultsof the average undrained shear strength of copper
tailings were compared. The sample with 90%
relative density had the smallest value of Su ave while
the largest has a 60% relative density. When theactual relative density was computed and comparedwith the target value, an error ranging from 0.66% to17.7% was observed. Since the samples werereconstituted, difficulties in achieving the sametarget relative density and soil structure at minimalerror can be encountered. It can also be observed

Table 4. Summary of void ratio (e) for CU test.
435

GEOMATE- Bangkok, Nov. 14-16, 2016

Dr

σ3

(%)

(kPa)

60

25

90

e targetslope of NCL or compression index (λ) and yintercept of the NCL line (Γ). The equations
presented in Fig. 4 can be used to predict the specificvolume for a certain mean normal effective stress orvice versa. The coefficient of determination for NCL
at 100 kPa and 50 kPa is 0.8386 and 0.8631,
respectively. Based on these values, the equation has
the capability of predicting values at almost 85%
accuracy. The slope can be used to compute for thesettlement parameter such as compression index(C c ) using equation [14]:e actual
InitialAfter
Saturation

After
Consolidation

1.13

1.09

1.16

1.09

50

1.13

1.22

1.37

0.94

100

1.13

1.19

1.59

1.18

50

0.98

0.97

0.89

0.99

100

0.98

0.99

0.78

0.97

200

0.98

1.03

1.35

1.01

=
l

Table 5. Critical angle of interal friction (φ crit ).
D r (%)

Copper Tailings

Triaxial TestCU

φ crit
25°

CD

-

60

φ’ crit27°

φ crit
21°

28°

-

90

φ’ crit22°
24°

Critical state parameters are obtained from the
CD test based on the concepts from the critical statemodel [13]. Results are tabulated at Table 6. The
isotropic normal compression lines (NCL) of thecopper tailings with 90% relative density arepresented at Fig. 4.

Specific Volume, v

3
y = -0.0009lnx + 2.5206, R² = 0.9014
y = -0.0011lnx + 2.164, R² = 0.8631
2
y = -0.0007lnx + 1.9832, R² = 0.8386
1.5
1
0.5

50kPa
100kPa200kPaSpecific Volume, v

0
1
10
100
Mean Normal Effective Stress,lnp' (kPa)1000

NCL of copper tailings having D r =Fig. 490% under CD test.
Void ratio (e) can be extracted from Fig. 4 since,
specific volume (v) is [13]:2.35252.352
2.35152.351
2.35052.35
2.34952.349
2.34852.348
2.3475NCL
CSL

0

v= 1− e(2)

Values computed, using Eq. 2, were less than
the typical values ranging from 0.1 to 0.8 [14]. From
this result, Copper tailings can be considered asslightly compressible. The slope of the reloading or
unloading line (κ) is zero since this step was not
simulated in the experiment. The critical state line
(CSL) of deviator stress (q) versus mean normal
effective stress (p’) has a slope M, Fig 6. It can beobserved that sample with 60% relative density has alarger value. This is due to the collapse the sample
experienced at 200 kPa for 90% relative density.
After establishing the critical state parameters, thespecific volume versus p’ was redrawn using theformulas from the critical state model, Fig. 5. The
area beyond NCL is the impossible state while thearea below is where the over-consolidated soil islocated. Points along NCL are normally consolidated
soil. The area in between NCL and CSL is wherecompression occurs. The area under CSL is where
expansion is encountered.

Critical-State Parameters

2.5

Cc
Cc
=
ln(10) 2.3

(1)

200

400

600

800

Mean Normal Effective Stress, p' (kPa)Specific volume versus mean normal
Fig. 5effective stress plot of copper tailings having D r =90% under CD test.

From the computations, the void ratio for 200
kPa NCL has values larger than e max . Exceeding
e max can be an indication that the sample had alreadycollapse at that confining pressure. This is also thereason of having a smaller φ crit or φ’ crit at 90%
relative density. The sample had already a weakerstructure prior to the shearing phase. With this, theNCL for 100 kPa is considered as its maximum
boundary while NCL for 50 kPa is the minimum
boundary. Each data was linearized to determine the

Cam Clay Model
The Cam Clay model was proposed by Roscoe
and Schofield (1963). It considers the yield behaviorof a soil through a yield locus as seen in Fig. 6. The
yield locus is defined as the combination of thedeviator stress, q, and the mean normal effective
436

GEOMATE- Bangkok, Nov. 14-16, 2016

stress, p’, that caused the yielding of a soil due to a
preconsolidation stress, p’ o . The preconsolidationstress is estimated from the deviator stress and mean
normal effective stress at critical state [13].
q
p'
+ ln(
)=
0
Mp '
p 'c

amount of fines content. Based on the AASHTO
rating it can still be as an embankment material butwith proper control. Based on the CU and CD test,
the effective critical angle of friction has a range of
21°-28°. The material performs best at 60% relative
density at critical state. When the critical stateparameters are incorporated to the Cam Clay model,
its maximum mean normal effective stress anddeviator stress is 812.91 kPa and 297.15 kPa
respectively.
The result presented in the research is notenough to completely assess Copper tailings’
potential as a construction material. California
Bearing Ratio test must be performed to further
evaluate its subgrade strength [17]. Furthermore, theCopper tailings’ performance as a fine aggregate
should be tested [18], [19].

(3)

The yield locus serves as the boundary of elastic
and plastic state [7]. Elastic state is within the locus
while plastic state is beyond it. The model wasapplied since the sample experienced yield behavior.
Furthermore, the model also integrates the criticalstate parameters together with NCL and CSL in its
formulation. The model can be used to predict thepossible deviator stress with a desired value of mean
normal effective stress or vice versa to assess the
behavior of the sample. The yield locus also shows
the limiting values for deviator stress and mean
normal effective stress.
Deviator Stress, q (kPa)

350

[1]

200kPa

Yield
Locus

300

REFERENCES

50kPa

[2]

100kPa

250

[3]

200

CSL

[4]

150
100

[5]

50
[6]

0
0

100 200 300 400 500 600 700 800 900[7]Mean Normal Effective Stress, p' (kPa)Stress path with yield locus of copper
Fig. 6tailings having D r = 90% under CD test.

[9]

Table 6. Summary of critical-state parameters[10]

Triaxial
Test
CopperTailings

CD[8]

D r (%)

λ
Γ
κ
M

[11]

60

90

0.00020.0005
1.71612.0176
0
1.13

0.00070.0011
1.98322.1640
0
0.99

[12]
[13]
[14]

CONCLUSION[15]

It is the interest of this study to assess thepossibility of using mine tailings, specifically copper
tailings, as a construction material. Index properties
and critical shear strength parameters were firstestablished following ASTM standards. From this, it
was established that the copper tailings can beclassified as silty soil with plasticity and a sufficient

[16]

[17]

[18]

437

Senate Economic Planning Office (2005). Retrieved fromhttps://www.senate.gov.ph/publications/PI%20200511%20%20Extracting%20Growth%20from%20Mining.pdf
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predicting hydrauliconductivity of mine tailings. International
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Wise Uranium Project (2002). Retrieved from http://www.wiseuranium.org/new02.htm
Vigar, A.J., Motton, N.T., & Taylor, I. (2011). Competent person’sreport on the Basay copperproject, Negros Oriental province,
Philippines. Hong Kong: Mining Associates.
Alshible, K., Batiste, S. & Sture, S. (2003). Strain Localization in
Sand: Plane Strain versus Triaxial Compression. Journal of
Geotechnical and Geoenvironmental Engineering. 129(6).483-494.
Liu, C & Evett, J.B. (2000). Soil Properties: Testing, Measurement,
and Evaluation. Upper Saddle River, New Jersey: Prentice-Hall,
Inc.
Britto, A.M. & Gunn, M.J. (1987). Critical State Soil Mechanics via
Finite Elements. West Sussex, England: Ellis Hordwood Limited.
Das, B. & Sobhan, K. (2014). Principles of GeotechnicalEngineering Sixth Edition, United States: Chris
Carson.
Qui, Y. & Sego, D. C. (2001). Laboratory Properties of Mine
Tailings. Canadian Geotechnical Journal. 38. 183-190.
Coduto, D. P. (1999). Geotechnical Engineering Principles andPractice. Upper
Saddle River, New Jersey: Prentice-Hall, Inc.
Shamsai, A., Bateni, S. & Ayatollahi,S. (2007). Geothecnical
Characteristics of Copper Mine Tailings: A Case Study. Springer
Science+Business Media. 25. 591-602.Marr WA, “Advances andRetreats in Geotechnical Measurements”, Judgment and Innovation:
The Heritage and Future of the Geotechnical Engineering Profession,
2000, pp. 42-60.
Smith, I., 2006 Smith’s Element of Soil Mechanics 8th Edition
Blackwell Publishing
Powrie, W. (2004). Soil Mechanics Concepts and Applications
Second Edition. London: Spoon Press.
Budhu, M. (2011). Soil Mechanics and Foundations. New York:John Wiley & Sons, Inc.
Bardet, J.P. (1997). Detained description of Triaxial Tests on CoarseGrained Soils.
Chapter 7.7 of Experimental Soils Mechanic.
Upper Saddle River, New Jersey: Prentice-Hall.
United States Department of Agriculture. (1994). Slope Stability
Reference Guide for National Forests in the United States.
Washington, DC. 2.
Prahallada, M. C. and Shanthappa, B. C. (2014). Use of Copper Ore
Tailings - as an Excellent Pozzolana in the Preparation of Concrete.
International Journal of Advanced Research in Engineering andApplied Sciences. 3(3). ISSN: 2278-6252
Thomas, B. S., Damare, A. and Gupta, R. C. (2013). Strength anddurability characteristics of copper tailing concrete. Construction and
Building Materials. 48, pp. 894-900.

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

GEOTECHNICAL PROPERTIES OF LADLE FURNACE SLAG IN
ROADWORK APPLICATIONS

Farshid Maghool1, Arul Arulrajah2, Suksun Horpibulsuk3 and Yan-Jun Du4
Department of Civil and Construction Engineering, Swinburne University of Technology, Australia;
3
School of Civil Engineering, Suranaree University of Technology, Thailand; 4 Institute of GeotechnicalEngineering, Southeast University, China
1,2

ABSTRACT
Ladle Furnace Slag (LFS) is an industrial waste from steelmaking process, generated during the final stage inladle refining furnace. LFS samples were sourced from a steelmaking company in Victoria, Australia. This paper
reports on the results of a geotechnical and chemical evaluation on LFS to assess the viability of using this byproduct in roadwork constructions. A comprehensive suits of engineering tests was undertaken comprisingmoisture content, specific gravity, particle size analysis, hydrometer, organic content, flakiness index, Los Angeles(LA) abrasion, hydraulic conductivity, modified compaction, pH, Unconfined Compression Strength (UCS) andCalifornia Bearing Ratio (CBR). In addition, the chemical composition of LFS sample was assessed using X-ray
Fluorescence analysis (XRF). The engineering properties of LFS were then compared with typical quarry materials.
The chemical composition results indicate that LFS contains high amount of calcium. Thus, the effect of 0, 7 and
28 days of curing on the strength of unbound LFS was investigated using UCS test. The specimens with 7 and28days curing had higher UCS value than uncured samples attributed to high lime content and hydration process.
From an engineering perspective, LFS was found to be suitable material to substitute typical quarry material inroadwork applications. The reuse of LFS in roadwork construction will significantly reduce the demand of natural
quarry aggregates and will possibly divert huge amount of these waste from landfills and stockpiles. Possible
applications of unbound LFS in road pavement are also discussed.
Keywords: Ladle Furnace Slag, Pavement Base, Unconfined Compression Strength, Industrial Waste
INTRODUCTION

molten refined steel is poured out of the furnace andthe resulting product on the bottom of the ladle is
known as LFS. On average the production of one
tonne of steel in steelmaking plants results in 30 to 50
kg of LFS [5].
Limited research on engineering and geotechnicalproperties of LFS have been conducted up to thepresent time. Serjun et al. [6] graded LFS as a low
quality material due to its fine grain size, adverse
leaching potential and expansive behavior. Manso etal. [7] indicated LFS can be a valuable by-productafter it has been turned into a dusty product and its
expansive characteristics could be reduced in thisform.
Manso et al. [8] studied the behavior of soils
stabilized with LFS and found similar results of thesame soils after mixing with lime. This research alsosuggested that UCS value was significantly increasedin the soils blended with LFS powder while theplasticity index and free swelling behavior of soils
can be reduced. Serjun [6] stated that LFS has thepotential to be used as a supplementary cementing
material in numerous civil and construction
applications due to its cementitious hydraulicproperties.

Waste materials are universally described as byproducts of Commercial, industrial, building anddemolition activities that do not have any value [1].
Due to the recent enforcement of more stringent
environmental regulations all around the world,
recycling and reuse of waste materials has become
very critical.
Steel has been known as the one of the world’s
most recyclable materials. Annually, more than 1400
million tonnes of steel is manufactured worldwide[2]. The steelmaking process generates an industrial
by-product termed as slag. In Australia alone, theannual production of steel slag is reported to bearound 2.4 million tonnes [3].
In the past decade, noticeable amount of slags was
effectively utilized as cementitious and noncementitious construction materials. More than 60%of the utilized slags in Australia was granulated blast
furnace slag, which is really popular among cement
and concrete companies [4]. However, there are
currently limited studies and reuse options forparticular types of slag including LFS.
LFS is formed during the secondary steelmaking
process in the ladle refining furnace. In this stage, the438GEOMATE- Bangkok, Nov. 14-16, 2016

The use of industrial wastes as roadwork
construction materials will be an optimistic outcome
for the waste management hierarchy, provided thatthe use takes into account the required environmental
and engineering considerations [9]. The limited
knowledge on the relevant engineering andgeotechnical properties of unbound LFS is main
obstacle for using these materials in roadwork
applications. An extensive suite of engineering andchemical testing were undertaken on LFS to evaluate
the feasibility of using these steel slag aggregates asroad construction materials.

CBR, UCS and hydraulic conductivity samples
were prepared and compacted under modified
compaction effort at optimum moisture content toreach at least 98% of maximum dry density. Threesets of sample were prepared for each of these tests.
The CBR samples were then soaked for 4 days with asurcharge mass of 4.5 kg on top. The swelling of the
CBR samples were measured during the 4 days ofsoaking using a dial gauge supported by a metal
tripod. UCS samples were compacted in five layers ofpredetermined mass in a spilt mold with an internal
diameter 100 mm and effective height of 200 mm. To
evaluate the effect of curing on mechanical propertiesof unbound LFS, two extra sets of UCS samples wereprepared and cured in a humid chamber with aminimum humidity of 95% and temperature controlof 25°C for 7 and 28 days.

MATERIALS AND METHODS
Samples for this research were collected from the
top of various LFS stockpiles from a steelmaking
plant in Victoria, Australia. LFS is well-graded grey
material and produced during the secondary
steelmaking process in the ladle refining furnace. LFS
was poured out from the ladle furnace in a liquid-state
and then is cooled down from 1600°C to room
temperature [7]. To obtain a representative sample forlaboratory testing, LFS samples were properly mixed,
split and sieved through a 20mm sieve.

Table 1 Chemical composition of LFS.
Chemical compounds
Fe2O3
CaO
SiO2
MgO
MnO2
Cr2O3
TiO2
SO3
P 2O 5K 2O
Al2O3

Weight (%)
35.23324.89922.934
8.633
5.827
0.954
0.498
0.498
0.466
0.059
non-detected

RESULTS AND DISCUSSION
Chemical composition results of LFS are reported
based on 100% normalization of oxide compounds
and presented in Table 1.

Fig. 1 LFS sample appearance.
Fig. 1 presents the LFS sample appearance after
passing through a 20 mm aperture sieve. It is evident
from the photo that LFS sample in this study contains
some electric arc furnace slag and natural aggregatesfrom the manufacturing site plant. The chemicalcomposition of LFS samples were assessed using Xray Fluorescence analysis (XRF). The natural
moisture content of LFS was also determined using
drying oven set at 105°C.
A series of laboratory evaluations was conducted
to assess the engineering properties of the LFS
comprising particle size distribution [10], hydrometer[11], organic content [12], flakiness index [13],particle density and water absorption [14]-[15], pH[16], modified compaction [17], LA abrasion [18],Hydraulic conductivity [19], CBR [20] and UCS [21],according to relevant established Australian andAmerican standards.

100

Percentage Passing (%)

90

LFS - Sieve analysis

8070

LFS - Hydrometer

60
50
40
30
20
10
0
0.001

0.01

0.1

1

Particle Size (mm)Fig. 2 Particle size distribution plot of LFS.
439

10

100

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 2 presents the engineering properties of
LFS sample. The initial moisture content of sample
was measured in a drying oven and found to be lessthan 2%.

LFS had a LA abrasion value of 31, which is in
the range of the values accepted by state roadauthority. Hydraulic conductivity of the LFS sample
was measured to be 9.81×10-09 m/s, which can belabelled as a very low permeable material inaccordance with the hydraulic conductivity
classification chart presented by Terzaghi et al. [24].
The modified Proctor compaction test results
indicated that the LFS had high maximum dry densitywith relevant optimum moisture content, higher thanthat of other recycled materials which have beenpreviously used in road applications [23]. The highmaximum dry density value can be attributed to very
high apparent particle density of the LFS aggregates.

Table 2 Engineering properties of LFS.
LFS
<2
0.5
13.5
48
38
SM
0.5
3.39
3.43
12.2
30
31
9.81×10-09
2.64
9.2

70
Test 160

Load on Piston, kN

Engineering parameters
Natural moisture content (%)
Clay content (%)
Silt content (%)
Sand content (%)
Gravel content (%)
USCS natation
Organic content (%)
Particle density: fine (kN/m3)
Particle density: coarse (kN/m3)
pH
Flakiness index (%)
LA abrasion loss (%)
Hydraulic conductivity (m/s)
Maximum dry density (Mg/m3)
Optimum moisture content (%)

The organic content value of LFS was found to bevery low, possibility due to the high temperaturesduring the steelmaking process. The both (fine andcoarse) apparent densities of LFS were very high,higher than that of typical quarry materials (>2
kN/m3) and very close to the values previously
reported for electric arc furnace slag [23]. The pH
values of samples indicate that LFS is alkaline bynature. The flakiness index of LFS was found by
separating the flaky particles and was noted to beapproximately 30.

Test 3

50
40
30
20
10
0
0

1

2

3

4

5

6

7

8

9

10 11 12

Penetration, mm

Fig. 3 CBR Load-penetration curve of LFS.
Table 3 present the strength properties of LFS
sample. Fig. 3 indicates the load penetration curves
obtained from three CBR tests on unbound LFS. The
CBR tests were performed after 4 days of soaking
condition where the samples believed to be saturated
and matric suction is insignificant. The CBR value
range of 156% to 165% indicates that the LFS meetsthe minimum requirement of 80% for typical quarrymaterials to use in pavement applications [23]. The
CBR values for the LFS is higher than of other wastematerials such as recycled concrete aggregate,crushed brick and reclaimed asphalt pavement used inroad applications [25]. The swell characteristic of the
LFS was measured during the soaking condition andit was found to be less than 0.5%.

Table 3 Strength properties of LFS.
Strength properties
CBR (%)
Swell (CBR) (%)
UCS_ no curing (MPa)
UCS_7 days curing (MPa)
UCS_28 days curing (MPa)

Test 2

LFS
156-165
<0.5
0.2
0.8
0.9

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GEOMATE- Bangkok, Nov. 14-16, 2016

The LFS was classified as high calcium material
due to high lime content, approximately 25% inchemical composition. The modified compaction test
indicates that LFS has high maximum dry density
value which can be attributed to high apparent
particle density. The LFS can be classified as very
low permeable material based on hydraulicconductivity test result.
The swelling of the LFS sample was measured
during the soaking condition of CBR and it was foundto be less than 0.5%. LFS was found to have muchhigh CBR value, higher than of typical quarrymaterials and other waste materials previously usedin base/subbase applications.
The UCS test results indicate that the UCS valueof LFS can significantly increase with curing time.
The significant increase in UCS can be attributed tohigh lime content and hydration process as chemicalbonding develops over time. The hydration reaction
was faster and more effective during the first 7 daysof curing.
In term of usage in roadworks application, theengineering properties of LFS were found to beequivalent or even superior to that of typical quarrymaterials. LFS can be suitable as a construction
material for pavement bases/subbase and engineering
fills. The reuse of LFS in roadwork applications will
significantly reduce the demand of virgin materials
and will potentially divert significant quantities ofthese by-products from landfills and stockpiles.

1.2

Shear Strength (MPa)

LFS - no curing
1

LFS - 7 days curing
LFS - 28 days curing
12.5%

0.8

0.6

300%

0.4

0.2

0

Fig. 4 Development of UCS in LFS with Curing.
The UCS values for unbound LFS with 0, 7 and
28 days of curing were evaluated and the effect of
curing time on the UCS value is shown in Fig. 4.
Three sets of samples were prepared for eachcondition and the average UCS value for eachcondition was individually reported in Table 3. The
UCS value of unbound LFS was found to increase
considerably with curing time. The significant
increase in UCS value was faster and more effective
during the first 7 days of curing, which can be
attributed to hydration process as chemical bonding
develops over time. The curing period was found toplay a vital role in the UCS development of unbound
LFS samples, which was attributed to the relatively
high CaO content of the LFS material.

ACKNOWLEDGEMENTS
The authors would like to thank Harsco Metals& Minerals Asia Pacific (Victoria, Australia) for
providing the materials in this study.

CONCLUSIONREFERENCES

The engineering and geotechnical laboratory
evaluation undertaken on LFS to assess the viabilityof using this by-product in roadwork constructions.
The LFS samples were sourced from a steelmaking
company in Victoria, Australia. The LFS wasclassified as silty sand (SM) and particle size
distribution plot indicates that the LFS has thepotential to be used as pavement materials. LFS was
found to be non-plastic with low cohesion and highsilt grain size content in the fine fractions.
The durability and LA abrasion loss of LFS
aggregates are found to be within the requirements oftypical quarry materials used in pavement
applications. Natural moisture content and organic
content of LFS was found to be very low. Both fineand coarse apparent density of LFS indicate theexistence of high quality material and aggregate in
LFS sample. The pH value of LFS was measured tobe 12.2, which indicates LFS is alkaline by nature.

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waste fact sheet, Publication No. 1440.
2012, Environmental Protection Agency ofVictoria, Australia.
AS, Methods for sampling and testing
aggregates Method 11: Particle size
distribution by sieving, in AustralianStandards AS-1141.11. 1996, Standards
Australia: NSW, Australia.
ASTM, Standard test method for particlesize analysis of soils. ASTM Standard D422.
1963,
ASTM
International:
West
Conshohocken, PA.
ASTM, Standard Test Methods for
Moisture, Ash, and Organic Matter of Peatand Other Organic Soils. ASTM Standard
D2974. 2007, ASTM International: WestConshohocken, PA.
BS, Method for Determination of Particle
Shape; Flakiness Index. British Standard
812-105.1.
2000,
BritishStandard
Institution: London, UK.
AS, Particle density and water absorption offine aggregate. Australian Standard 1141.5.
2000, Australian Standard: Sydney,
Australia.
AS, Particle density and water absorption ofcoarse aggregate - Weighing-in-water
method. Australian Standard 1141.6.1.
2000, Australian Standard: Sydney,
Australia.
AS, Soil chemical tests—Determination of
the pH value of a soil—Electrometricmethod. Australian Standard 1289.4.3.1.
1997, Australian standard: Sydney,
Australia.

AS, Soil compaction and density tests Determination of the dry density/moisture

[19].

[20].

[21].

[22].

[23].

shear strength responses of recycledconstruction and demolition materials in
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base/subbase
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Materials, 2014. 58: p. 245-257.

442

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

TEMPERATURE EFFECTS ON ELASTIC STIFFNESSOF HDPE GEOGRID AND ITS MODELLING
Thitapan Chantachot1, Warat Kongkitkul2 and Fumio Tatsuoka3
1,2
Department of Civil Engineering, Faculty of Engineering,
King Mongkut’s University of Technology Thonburi, Thailand;
3
Department of Civil Engineering, Tokyo University of Science, Japan

ABSTRACTThe ambient temperature is one of the important factors that affect the mechanical properties of polymer
geosynthetic reinforcement. With an increase in the temperature, the rupture strength and the elastic stiffness
decrease. In this study, to understand the temperature effects on the load-strain-time behaviours of a polymer
geogrid, a series of tensile loading tests were performed on a high-density polyethylene (HDPE) geogrid atdifferent but constant temperatures, and also under step-increasing temperature conditions. The test results
indicated that the elastic stiffness of tested geogrid increases with the load level, while decreases with the ambient
temperature. These properties were modelled based on the framework of hypo-elasticity. An existing non-linear
three-component (NTC) model, which can simulate the load-strain-time behaviours of many types of polymer
geogrid subjected to arbitrary loading histories (e.g., monotonic loading at different rates, creep, load relaxation)
under a constant temperature, was modified to account for the dependency of the elastic stiffness on the load level
and the temperature, as well as the dependency of the rupture strength on the temperature. The modified model
can simulate very well the observed temperature effects on the elasticity of the tested geogrid.
Keywords: Elastic stiffness, Geogrid, Non-linear three-component model, Temperaturetemperature [5]-[8]. Therefore, the study oftemperature effects on the load-strain-time
behaviours of geosynthetic reinforcements isimportant.
To simulate the rate-dependent behaviours ofgeosynthetic reinforcements (i.e., monotonic loading
at different but constant strain rates, creep, and load
relaxation), an elasto-viscoplastic non-linear threecomponent (NTC) model was proposed [9].
Subsequently, the NTC model was modified tosimulate temperature-accelerated creep tests by
taking into account the effects of temperature increaseon the rupture tensile strength and stiffness ofgeosynthetic reinforcement [10]. The simulation
results indicated that the NTC model is capable ofsimulating both rate- and temperature-dependent
behaviours. However, in this previous study, theelastic stiffness was treated unaffected by thetemperature rise, and validation of the model withexperimental data has not been performed.
In view of the above, in this study, a series oftensile loading tests were performed to evaluate
effects of temperature and load/temperature history
on the elastic stiffness of a HDPE geogrid. Then, theNTC model was modified to realistically simulate thetest results by incorporating the dependency of theelastic stiffness on the temperature.

INTRODUCTION
A number of different types of polymer
geosynthetic
reinforcement
are
used
as
reinforcements for various geotechnical engineering
structures (e.g., slope, embankment, and retainingwall). Typically, constructions of these structuresutilising geosynthetic reinforcements are faster andmore cost-effective than ordinary conventional
methods. In design of a geosynthetic-reinforced soil
(GRS) structure, it is required to know the mechanicalproperties of a given geosynthetic reinforcement,
especially the rupture tensile strength. However, therupture tensile strength is usually determined by anambient temperature condition that is controlled
constant throughout a test (equal to 20 oC following
ASTM D4595 and EN ISO 10319) [1]-[2].
On the other hand, air-temperature cyclicallychanges both daily and seasonally. This variation of
air-temperature affects the ambient temperature of
backfill in a GRS structure, and thereafter, thegeosynthetic reinforcements arranged inside [3].
From field monitoring of a GRS structure, changes of
air-temperature were found to correlate with thegeogrid strains arranged inside [3]-[4]. Moreover,effects of temperature rise on the load-strain-time
behaviours of geosynthetic reinforcements were alsostudied in the laboratory. It was found that not only
the rupture strength but also the pre-peak stiffness andthe elastic stiffness decrease with an increase in the443

GEOMATE- Bangkok, Nov. 14-16, 2016

ELASTIC STIFFNESS AS A FUNCTION OF
LOAD LEVEL AND TEMPERATURE

80
70

Monotonic Loading (ML)60

Tensile load, V (kN/m)

To determine the elastic stiffness, k eq , of ageosynthetic reinforcement, small-amplitude unloadreload cyclic loading tests have been performed on
many types of geogrid [7]. Then, the following
formulations of dependency of k eq on the load level
and the temperature was proposed:

o

30 C

50
o

40

35 C
o

50 C
30

o

45 C
o

40 C

20
10

 : End of test
: Tensile rupture strength

0

 V 
k eq  k eq0 (T )  Vmax  

k eq0  p ' Af 

m
0

5

(1)

10

15

20

25

30

35

Tensile strain,  (%)

q'

(2)

Fig. 1

where: keq0 (T ) is the value of k eq when V / Vmax = 1;

Tensile load-tensile strain relations fromcontinuous ML tests under different constanttemperatures

Table 1 Rupture strength under different constanttemperatures

V / Vmax is the load level, defined as ratio of the tensileload, V , to the maximum tensile load at respective
ambient temperature, Vmax ; m , p ' and q ' are thematerial constants, depending on types of polymer

Temperature (oC)
Rupture strength (kN/m)
30
53.2
35
48.1
40
44.9
45
41.8
50
38.9
Note: rupture strength was defined for a strain rate at
rupture equal to 0.1 %/min

geosynthetic reinforcement; and Af  is thetemperature effect parameter, defined as ratio of the
rupture strength at a given temperature to the rupture
strength at standard temperature of 20 oC. Note that
rupture tensile strength is defined for a strain rate at
rupture equal to 0.1 %/min.
TEST MATERIAL AND PROGRAMME

these tests, any rupture was not observed at the ends
of test (denoted by symbol “+”), due to a limited
maximum stroke of the tensile loading machine.
Therefore, the rupture tensile strengths were definedat the points of maximum curvature along therespective V -  relations (denoted by symbol “×”).
Then, these values were corrected to the values at thesame strain rate (0.1 %/min) to eliminate the effectsof strain rate on the tensile rupture strength ofgeosynthetics [7]. The values of corrected rupture
strength are summarised in Table 1. It can be seen thatwith the increasing temperature, the rupture strength
decreases.
Figure 2 shows a relationship between the

In this study, to determine the elastic stiffness andthe temperature effect parameter, a series of special
tensile loading tests were performed on a high-density
polyethylene (HDPE) geogrid, employing thefollowing two different loading patterns: i)continuous monotonic loading (ML) until the rupture;and ii) sustained (creep) loading (SL) applied duringotherwise ML until rupture. A constant load rate of
0.6 kN/m/min was used for ML in both loading
patterns i) and ii). In loading pattern ii), ML was first
performed until achieving the specified tensile load
and then SL was held for three hours, after whichanother ML was restarted until the rupture. The
ambient temperature surrounding the test specimen
was controlled constant throughout a test at either 30,
40 or 50 oC by means of a temperature-controlledchamber [7].

temperature effect parameter, Af  , and temperature.
The value of Af  is unity at the standard temperature,
T0 , of 20 oC. The procedures to determine the A 
fvalue are explained in details in Kongkitkul et al.

TEST RESULTS AND DISCUSSIONS(2012) [7]. It can be seen that the Af  value decreases
from unity at T0 = 20 oC to a smaller value at a higher
temperature. Thus the reduction of rupture tensile
strength with the increasing temperature of HDPE
geogrid can be described by Eq. (2), which

Temperature Effect Parameter
The relationships between tensile load, V , andtensile strain,  , obtained from ML tests at different
but constant temperatures are shown in Fig. 1. In

incorporates Af  .

444

GEOMATE- Bangkok, Nov. 14-16, 2016

80

1.1

o

Temperature = 30 C
70

Tensile load, V (kN/m)

1.0

A

f'

0.9

0.8

0.7

f'

A = Vmax/V max0 = 1 - a (T-T'0)/T'0)f'

A = 1-0.26149((T-T'0)/T'0)0.6

b0.90059

 : End of test
: Tensile rupture strength

60

ML
50
40

74 %
30

62 %
49 %

20

37 %

10

2

R = 0.98766

SL (load level, V/Vmax = 25 %)

0
0.5
15

20

25

30

35

40

45

50

55

Remark: Vmax@30 C = 53 kN/m
o

0

5

15

20

25

30

35

Tensile strain,  (%)

oTemperature, T ( C)a)Temperature effect parameter, Af  , as afunction of temperature

80oTemperature = 40 C
70

Tensile load, V (kN/m)

Fig. 210

Elastic Stiffness
Determination of elastic stiffness
Figures 3(a) to 3(c) show the V -  relations of
the HDPE geogrid by SL under different temperaturesequal to 30, 40, and 50 oC, respectively. In thesefigures, SL tests were performed at different tensile
load levels, V / Vmax . The values of Vmax at respective
temperature are shown in Table 1. It can be seen thatV -  relation of the geogrid exhibits a very highstiffness, close to the elastic value when ML wasrestarted following SL [9]. Therefore, in this study,
the elastic stiffness was determined from a linear
relation fitted to the initial V -  curve after therestart of ML following SL, as shown in Fig. 4.

 : End of test
: Tensile rupture strength

60
50

ML

40

88 %
30

59 %

20
10

SL (load level, V/Vmax = 29 %)0

Remark: Vmax@40 C = 45 kN/m
o

0

5

10

15

20

25

30

35

Tensile strain,  (%)

b)
80
o

Temperature = 50 C

Tensile load, V (kN/m)

70

Effects of load level and temperature on elastic
stiffness
Figure 5(a) shows the elastic stiffness, k eq , versus

 : End of test
: Tensile rupture strength

60

ML

50
40
30

84%
67%

20

50%
10

load level, V / Vmax , relations in the full-log plot of
HPDE geogrid for different temperatures. The linear
lines fitted to the data points for respective
temperatures represent Eq. (1). It is readily seen thatk eq increases significantly with an increase in the

SL (load level, V/Vmax = 34 %)Remark: Vmax@50 C = 39 kN/m

0

o

0

5

10

15

20

25

30

35

Tensile strain,  (%)c)
Fig. 3load level, while drastically decreases with anincrease in the temperature. These behaviours
indicate that the k eq values of HDPE geogrid exhibit
a combination of the hypo-elasticity and thedegradation by temperature effects.
The behaviours of k eq determined by the method

Tensile load-tensile strain relations from SL
tests during otherwise ML under different
temperatures of: (a) 30 oC; (b) 40 oC; and (c)50 oCused to best fit to the k eq - V / Vmax data points
obtained in this study (Fig. 5(a)). It can be seen thatthe linear lines fitted to the data points from thepresent study are consistent very well with the data
points from Kongkitkul et al. (2012) [7]. In Fig. 5(b),the k eq0 values according to Eq. (1) obtained from theused in the present study are similar to those reported
by Kongkitkul et al. (2012) [7], in which the k eq
values of the same HPDE geogrid were determinedby measuring the slope of V -  relations during
small-amplitude cyclic loadings. Equation (1) was

fitted relations shown in Fig. 5(a) were plotted against

445

GEOMATE- Bangkok, Nov. 14-16, 2016

60

4000
3500

o

Temperature = 30 C

Elastic stiffness, keq (kN/m)

Tensile load, V (kN/m)

3000

50

ML

Elastic stiffness,
keq= 1889.27 kN/m

40

SL (Creep level = 60 %)
30
5

10

15

2000
1500

1000

m

: keq=keq0(T)*(V/Vmax) ; m=0.14
Remark: Vmax at the respective temperature

500
0.2

20

Tensile strain,  (%)

Fig. 42500o

T ( C) This study Kongkitkul et al. (2012)
30
40
50

0.4

a)

0.8

1

0.8

1.0

1.2

3200

Determination of elastic stiffness from V  relation immediately at the restart of ML
following SL

2800

This study:
Kongkitkul et al. (2012) :

2400
2000

keqo (kN/m)

the Af  values for the respective temperatures
obtained from Fig. 2. Then, Eq. (2) was fitted to theplotted k eq0 and Af  data from the present study. It

1600

This study:
f' q'

1200

f' 2.65

keqo = p'(A ) = 2955(A )2

R = 0.99777

800

may be seen that the fitted curve is also in good
agreement with the data points from Kongkitkul et al.
(2012) [7].

400
0
0.0

MODELLING OF ELASTIC PROPERTIES

0.2

Fig. 5For realistic simulation of the load-strain-time
behaviours of polymer geosynthetic reinforcements
under a constant temperature, an elasto-viscoplastic
non-linear three-component (NTC) model wasdeveloped [9]. This NTC model consists of elastic,
inviscid, and viscous components. In the elastic
component, elastic strain rate,  e , is determined byan hypo-elastic model having the elastic modulus,keq (V ) , that is a function of instantaneous tensile load,

0.4

0.6
f'

A

b)

Non-linear Three-component Model

Elastic stiffness as a function of load level
and temperature: (a) keq - V / Vmax relations;and (b) k eq0 - Af  relation

Simulation Results
The V -  relations from continuous ML tests at
different constant temperature are presented in Fig.
7(a), while those from tests in which SL was
performed during otherwise ML (as those shown in
Fig. 7(a)) are presented in Fig. 8 (a). Figures 7(a) and8(a) compare these experimental results and theirsimulations. It can be seen that the modified NTC
model can simulate very well the measured V - 
relations for various load/temperature histories. The
dependency of keq on the load level and temperatureV: e  V / keq (V )0.6

Load level, V/Vmax(3)

On the other hand, with an increase in thetemperature, T , the keq (V ) value at a given load

can be seen very well from the tensile load, V , versus
the elastic tensile strain,  e , relations shown in Figs.
7(b) and 8(b). The tangential slope of V -  e relation
is equal to the current keq value. It can be seen fromlevel decreases (Fig. 5(a)). To take into account the
coupled effects of load level, V / Vmax , andtemperature, T , on elastic stiffness, keq (V ,T ) , theNTC model was modified as follows. Figure 6 shows
the method to obtain the current elastic stiffness fromthe current temperature, T , the current tensile load V, 0.
and a known rupture strength at T0 = 20 oC, Vmax
By implementing the algorithm shown in Fig. 6 into
the modified NTC model, the realistic current elastic
stiffness can be easily determined.

Fig. 7(b) that, at any given V value, the keq value
decreases with an increase in the temperature. On theother hand, at the same temperature, the keq value
increases with an increase in V . This result indicatesthat the modified NTC model can successfully
simulate the coupled effects of load level andtemperature on the elastic stiffness.
446

GEOMATE- Bangkok, Nov. 14-16, 2016T
= 20 oC (Fig. 2):

Eq (2):(Fig. 5(b))

Eq (1):(Fig. 5(a))

ML tests at
different
temperatures

SL tests at different
load levels and temperatures

Fig. 6 Diagram to obtain the current elastic stiffness value (modified from Kongkitkul et al., 2012)
The experimental and simulation results describedabove are obtained under the conditions of constanttemperature. To further examine the capability of themodified NTC model, a series of tests in which thetemperature was increased during SL from 30 oC at
the start of SL to 50 oC at the end of SL was performed.
Then, the results of these tests and their simulations
are presented and compared in Fig. 8(a). Figure 8(b)compares the simulated V -  e relation for the SL testin which the temperature was increased from 30 oC to
50 oC during otherwise ML with the simulated V -  erelations in continuous ML tests during which a SL
was performed at an intermediate load level under
different constant temperatures of 30, 40, and 50 oC.
It can be readily seen that the V -  e states during SL
tests under either constant or varying temperature do
not move as the tensile loads are kept constant. As aresult, the entire V -  e relation from the origin to theend of an intermission of ML by SL under theconstant temperature condition does not change likeobtained by a continuous ML test at a constanttemperature. On the other hand, upon the restart ofcontinuous ML at an elevated temperature (i.e., 50oC) after a SL test during which the temperature
increases from 30 oC to 50 oC, the V -  e relation
starts from a point along the V -  e relation from thecontinuous ML at a constant temperature of 30 oC
while exhibiting a drastic decrease in the slope to theone of the V -  e relation from the continuous ML ata constant temperature of 50 oC, as shown in Fig. 8(c).

8070Monotonic Loading (ML)o

30 C

Tensile load, V (kN/m)

60

o

40 C

50

o

50 C
40
30
20
V-

10

RuptureExperiment:
Simulation:

0
0

5

10



15

20

25

Tensile strain,  (%)

a)
80
70

Monotonic Loading (ML)Tensile load, V (kN/m)

60
50
o

30 C
40
o

35 C

30

o

40 C

20

o

45 C

10

o

50 C

0
0

1

2

3

4

5

6

7

e

b)
Fig. 7

CONCLUSIONS
The following conclusions can be derived fromthe experimental and simulation results that observedin this study:
1. Rupture tensile strength of the tested HDPE
geogrid decreases with an increase in the temperature.
2. Elastic stiffness increases with an increase in
the load level, while decreases with an increase in the

Elastic tensile strain,  (%)

Behaviours in continuous ML tests under
different temperatures: (a) measured andsimulated V -  relations; and (b) simulated
V -  e relations

temperature. These trends of behaviour can be
expressed by the coupled functional forms proposedin this study.
3. The modified NTC model can well-simulate
the variation of elastic stiffness, as a result of thecoupled effects between the load level and thetemperature
447

GEOMATE- Bangkok, Nov. 14-16, 2016

ACKNOWLEDGEMENTS
80

Sustained Loading (SL)70

The authors are grateful to the National ResearchCouncil of Thailand (NRCT), the Commission on
Higher Education, the Thailand Research Fund (TRF)
and the National Research University (NRU) Projectfor financial support. Thanks are also extended toVigor Merger Co., Ltd who provided geogrids for thisstudy.

(Creep level = 40 % of Vend of test, 30 C)oTensile load, V (kN/m)

60
50
o

40 C
40

o

30 C
o

30-50 C

30

o

50 C

20

REFERENCES

10
Experiment:
Simulation:

0
0

5

10

15

20

[1] ASTM International, “ASTM D4595-Standard
Test Method for Tensile Properties ofGeotextiles by the Wide-Width Strip Method”,
2011.
[2] International Organization for Standardization,“EN ISO 10319-Geosynthetics-Wide-Width
Tensile Test”, 2008.
[3] Fannin RJ, “Long-term variations of force andstrain in a steep geogrid-reinforced soil slope”, J.
of Geosynthetics International, Vol. 8, Aug.
2001, pp. 81-96.
[4] Tatsuoka F, Tateyama M, Koda M, Kojima K,
Yonezawa T, Shindo Y, and Tamai S, “Researchand construction of geosynthetic-reinforced soil
integral bridges”, accepted, 2016.
[5] Zornberg JG, Byler BB, and Knudsen JW,“Creep of geotextiles using Time-Temperature
Superposition methods”, J. of ASCE, Vol. 130,
Nov. 2004, pp. 1158-1168.
[6] Hsieh C, and Tseng YC, “Tensile creep behaviorof a PVC coated polyester geogrid at different
temperatures, J. of GeoEngineering, Vol. 3, Dec.
2008, pp. 113-119.
[7] Kongkitkul W, Tabsombut W, Jaturapitakkul C,
and Tatsuoka F, “Effects of temperature on therupture strength and elastic stiffness of geogrids”,
J. of Geosynthetics International, Vol. 19, Apr.
2012, pp. 106-123.
[8] Stępień S, and Szymański A, “Influence of strain
rate on tensile strength of woven geotextile in theselected range of temperature”, Studia
Geotechnica et Mechanica, Vol. 37, Dec. 2015,
pp. 57-60.
[9] Hirakawa D, Kongkitkul W, Tatsuoka F, andUchimura T, “Time-dependent stress-strain
behaviour due to viscous properties of geogrid
reinforcement”, J. of Geosynthetics International,Vol. 10, Dec. 2003, pp. 176-199.
[10] Kongkitkul W and Tatsuoka F, “A theoreticalframework to analyse the behaviour of polymer
geosynthetic reinforcement in temperatureaccelerated creep tests”, J. of Geosynthetics
International, Vol. 14, Feb. 2007, pp. 23-38.

25

Tensile strain,  (%)a)
80

Sustained Loading (SL)70

(Creep level = 40 % of Vend of test, 30 C)

Tensile load, V (kN/m)

o

60
o

30 C
50
o

40

30-50 C
o

30

50 C



SL stage20
10oRemark: "+" indicates the point
where the temperature was

40 C

0

o

o

increased from 30 C to 50 C
0

1

2

3

4

5

6

7

e

Elastic tensile strain,  (%)b)
60

Sustained Loading (SL)(Creep level = 40 % of Vend of test, 30 C)

Tensile load, V (kN/m)

50

o

o

30 C
40
o

30-50 C30
SL stage20ookeq(50 C)

keq(50 C)



o

50 C

o

keq(30 C)

Remark: "+" indicates the point
where the temperature was

10

o

o

increased from 30 C to 50 C
0
0

1

2

3

4

5

6

e

Elastic tensile strain,  (%)c)
Fig. 8Behaviours in continuous ML tests withintermediateSL
under
different
temperatures: (a) measured and simulated
V -  relations; (b) simulated V -  erelations; and (c) zoom-up of Fig. 8(b)448Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EVALUATION OF DIFFERENT FIBER REINFORCED MORTAR AS
RETROFITTING MATERIALS FOR RC COLUMNS
Bernardo A. Lejano1
Faculty, De La Salle University, Philippines

1

ABSTRACTThis study is about the use of fiber reinforced mortar (FRM) as a jacketing material for reinforced concrete(RC) columns. Presented in this paper are analyses of the test results of destructive-loading that were conducted
to explore the applicability of different jacketing materials as retrofit to RC columns. The jackets used to wrap
RC columns were built of mortar mixed with different fibers. The fibers considered were steel fibers andpolymer fibers. A simple model based on material properties was used to simulate the axial strength of thecolumns. The numerical model calculation results as well as the experimental test results were used to assess theperformance of the jacketing materials used. The experimental part consisted of testing two sets of column
specimens. The first set is wrapped with mortar jackets reinforced with steel fibers and the second set is wrappedwith mortar jackets reinforced with polymer fibers with and without fly ash. The efficacy of the different FRM
jackets as retrofitting materials was assessed by calculating the effect of confinement. The calculation was
accomplished by measuring the maximum load of the retrofitted columns with jackets and then subtracting theload that can be carried by the control specimens, that is, the column with the original cross section. The
confining effect was used as the basis in determining the confined compressive strength of concrete and thenused to establish a model that can simulate the axial strength of the columns. The results indicate that the FRM
jackets are effective as retrofitting material for RC column.
Keywords: Fiber Reinforced Mortar, RC Column, Retrofitting, Confining Effect, Axial Strength, Jacketing

to install. However, it is still very expensive.
As another option steel plates and FRP sheets,
jackets made of fiber reinforced mortar (FRM) areinvestigated in this study. This idea complements
other studies [4]-[6] that fiber reinforcement canincrease the strength, ductility and toughness ofconcrete. Even the use of recycled carpet fibers havebeen used as fiber reinforcement to concrete [7].
The study of Valerio et al. [8] recommended
Steel Fiber Reinforced Mortar (SFRM) as anotheroption of retrofitting material for RC columns. Theirstudy showed improvement in the load that can be
carried by the RC column after it is retrofitted with
SFRM. However, they reported that using steel
fibers has a down side. Since the fibers were stiffand cannot bend, it is difficult to place and distributethem in the column corners. This resulted to failureof the retrofitted columns initiated by the
propagation of cracks at the column corners.
Considering the problem stated above, analternative to steel fibers was proposed by Oropel etal. [9]. Polymer Fiber Reinforced Mortar (PFRM)
was used as jacketing material instead of steel fibers
because they are more flexible and can be easily
placed at the column corners. Moreover, aside frombeing light, these fibers are resistant to corrosion.
Two types of polymer fibers were used, synthetic
and cellulose fibers.
This paper focuses on the simulation of the load-

INTRODUCTION
Columns are primarily structural members of abuilding that should not fail because failure of evenone of these columns could result to the collapse of
the whole structure. When columns show some signof weakness, retrofitting is needed. Example to thisare aged reinforced concrete (RC) columns, whichusually are in need of retrofitting so as tocompensate for the lost strength due to various
environmental factors. Other reasons why RC
columnsneedrehabilitationareconcrete
deterioration, unexpected increase in load, or
physical damage. Retrofitting may be done byencasing the column with certain material. This
retrofitting technique is termed as jacketing. This is
the usual solution to fix deteriorated or damaged
columns caused by environmental hazards like fire,
seismic event, or differential upheaval of thefoundation because of expansive clay soil [1].
Jacketing using steel plates is the usual methodof retrofitting columns. While it is popular becauseof its simplicity, efficiency, and effectiveness, it also
has its criticisms. Aside from being expensive,
another disadvantage is that it may produce larger
earthquake forces because it is heavy. Studies [2],[3] have shown that fiber reinforced polymer (FRP)
sheet is a good alternative to steel plates. This
material is very light, does not corrode and very easy449

GEOMATE- Bangkok, Nov. 14-16, 2016

carrying capacity of the above-mentioned jacketing
materials (SFRM and PFRM). Previous work of theauthor [10] reported test results on the use of similar
retrofitting materials but lacks the model to simulate
the behavior of RC columns retrofitted with jackets.
This paper attempts to provide a simple model thatcan predict the axial strength of RC columns
retrofitted with FRM jackets once the confinement
effect is determined.

To make the mortar jacket suitable as retrofitting
material, it must possess the following qualities: low
drying shrinkage, fast strength development, andexcellent crack resistance. These qualities of mortar
may be attained by reinforcing it with fibers.
Moreover, shorter fiber reinforcement may bedesirable because it can be made into a thinner layerof repair [6]. However, the above mentioned
qualities of mortar jacket must be maintained.
One of the materials examined in this study is the
Steel Fiber Reinforced Mortar (SFRM) jacket.
Mortar was mixed with steel fibers to increase its
tensile strength. The steel fibers used in this study
have the following properties: unit weight is 19g,
length is 50mm and diameter is 0.75mm. Photos ofthe steel fibers are shown in Fig. 1.

CONCEPTS OF RETROFITTING USING FRM
The usual repair of RC column is done byremoving the damaged or deteriorated concrete andsubstituting it with new concrete. To improve its
strength, it may be encased with layer (or jacket) of
mortar or shotcrete [10]. The effect of the jacket is
two-fold, it enlarges the cross-section of the column
and it provides confinement to the original coreconcrete. In this study, the load is applied only tothe original section such that the effect of theenlargement of the section may not apply. Hence,
only the confinement effect will be emphasized inthis study. However, it may not be denied that thesection enlargement may have an effect because of
the redistribution of stresses.

Fig.1 Picture of steel fibers (left) and comparison ofa steel fiber with Phil. 5 peso coin (right)

Confinement Effect

The other material examined in this study is thePolymer Fiber Reinforced Mortar (PFRM). The twopolymer fibers used were the synthetic fibers and thecellulose fibers. These fibers were expected toprovide the same increase in tensile strength of
concrete as steel fibers provided, but more flexible.
The synthetic fibers are especially engineered
copolymer fibers designed as substitute to the steel
reinforcement in concrete floors on grade andpavements. They are also used as reinforcement inprecast concrete products with strength up to 35
MPa. As can be seen in Fig. 2, the size and shape ofsynthetic fibers suggest that it can be used as
substitute for steel fibers. The dosage used is 2.5 kg
per cubic meter of mortar [9].

When a material is subjected to axial
compression, it would be accompanied by transverse
tensile deformation. If this transverse deformation isrestrained, it would result to higher axial resistanceof the material. This phenomenon is the same forconcrete and is termed as the confinement effect. In
RC columns, the steel ties or hoops usually provide
the partial restrain of the transverse deformation
resulting to the confinement of the core concrete.
This usually results to an increase in the axial
strength of the column. In addition, the hoops avert
the concrete expansion and prevent tensile cracking
in concrete. Hence, considerable improvement inductility is usually observed [3].
In retrofitted RC columns, the jackets provide thesame confinement effect as mentioned above. To
provide adequate confinement the jacket must haveenough tensile strength. For this reason, the mortar
jacket is reinforced with fibers.
Material Selection
The choice of suitable material is an importantfactor in deciding the retrofitting technique to be
used. The material must match with the RC column
and must be easy to install. Prospective clients for
retrofitting jobs would usually want something thatis economical and that would not drastically impair
the use or occupancy of the structure. Reinforced
mortar jacket is an ideal material for this job.

Fig. 2 Pictures of synthetic fibers (left) and cellulose
fibers (right)
As shown in Fig. 2, the cellulose fibers appear
first as blocks but disperse evenly into concrete as
micro fibers after contact with water. Approximately
1500 million crack-fighting micro fibers are450

GEOMATE- Bangkok, Nov. 14-16, 2016

produced per kg of concrete on a dosage of 2 bags
per cubic meter of concrete. These cellulose fibers
have excellent bond with cement resulting todecrease in plastic shrinkage cracking. They also
tend to make the concrete denser. The same effects
are expected in mortar, hence improving the strengthof mortar jacket [9]. In addition, the cellulose fiber isalkali resistant and allows better surface finishing.

RC columns with square cross-section. The detailsof the dimension and reinforcement are shown in Fig.
3 and Fig. 4. Although the details shown are forcolumn specimens of PFRM series, the other series
were constructed in similar manner with slightvariation only in the size of the cross section and thejacket thickness. Furthermore, the ties are the samefor all column specimens justifying the assumptionthat the magnitude of its confinement effect (if there
is any) is the same for all column specimens.

EXPERIMENTAL PROGRAM
Several RC column specimens were made andsubjected to destructive testing to evaluate theirefficacy as retrofitting materials. In this study, 4cases were studied for the columns with SFRM
jacket (referred to as SFRM series). This includes
the control specimens (without jacket), andspecimens with mortar jackets containing 1%, 2%,
and 3% steel fibers by volume. This resulted to 3
control column specimens and 9 column specimens
with SFRM jacket. (Refer to Table 1). Sakthivel etal. [4] investigated concrete with almost the same
steel fiber volume range.

Fig. 3 Typical cross-section of column specimen
with PFRM jacket

Table 1 Construction details of specimens

Retrofitting
material
Original
column size
Concrete cover
Main rebars
Tie diameter
Tie spacing
Mortar jacket
thickness
Different cases(Each case has
3 column
specimens)

SFRM Series
Mortar w/
Steel Fiber
120x120x300
Ac=14400mm2
20 mm
4 – 10 mmφ
8 mm
160 mm
20mm

PFRM Series
Mortar w/
Polymer Fiber
180x180x400
Ac=32400mm2
40 mm
4 – 10 mmφ
8 mm
160 mm10mm1) control
2) 1% SFRM
3) 2% SFRM
4) 3% SFRM

1) control2) mortar
3) Synthetic4) Cellulous
5) Syn. w/ FA
6) Cel. w/ FA

Fig. 4 Typical side-section of column specimens
with PFRM jacket

There were 6 cases that were studied for columns
with polymer FRM jackets (referred to as PFRM
series). The 1st case is the control specimens
without jacket, the 2nd is with mortar jacket, the 3rd iswith synthetic FRM jacket, the 4th case is withcellulose FRM jacket, 5th is with synthetic FRM
jacket mixed with fly ash (FA), and the 6th case iswith cellulose FRM jacket mixed with fly ash. This
resulted to 12 column specimens with PFRM jacket,
3 column specimens with mortar jacket, and 3
control specimen (without jacket). (Refer to Table 1)
Shown in Table 1 are the construction details of
the RC column specimens. All columns were tied

Axial load was applied through steel plates, as
shown in Fig. 4, to ensure that the force is applied
only to the original cross-section of the column
specimen. This is done to simulate the possible
actual condition at the ends of the columns. Due todifficulties in application of the mortar at the column
ends, the jacket cannot be ensured to be in fullcontact with the other connected members of thestructure at the column ends resulting to ineffective
transmission of the axial load through the jacket
cross-section.

451

GEOMATE- Bangkok, Nov. 14-16, 2016

the maximum increase is 18.5%, which was obtained
at 2% steel fiber reinforcement.
It is interesting to note that there is no significant
increase in P max when plain mortar jacket was used.
This means that enlargement of the section due tomortar jacket did not contribute to the strength
increase of the column. Also, it seems that the ties
did not provide significant confinement. This was
also observed by other research [11] when the tie
spacing is medium to large. Furthermore, theintroduction of fly ash to the mortar mix did not
result to a better jacket.
Shown in Fig.5 are pictures showing thedifference in the crack patterns between column withPFRM jacket and column with SFRM jacket, at the
final failure. It can be seen that cracks are more
scattered in the PFRM series while cracks areconcentrated in the corner of the column with SFRM
jacket. Since it is difficult to place the steel fibers in
the corners of the SFRM jacket, the corners may beconsidered as zone of weakness.

TEST RESULTS
The results of strength tests of concrete, mortar
with fibers, and steel reinforcing bars are tabulated
in Table 2. It can be seen that the compressive
strength of mortar increased with the addition offiber. However, the introduction of fly ash in the
PFRM series resulted to the decrease in strength.
Fly ash was introduced because it was thought toimprove the performance of mortar. However, theresult was contrary as indicated in the mortar
strength test results shown in Table 2. It is not clear
why the fly ash did not perform as expected, but it
may have something to do with its proportion to themix. In this study, fly ash was used as 30%
replacement to cement.
Table 2 Strength of concrete, mortar, and steel
Material
Concrete,fc’
Mortarfc’ (MPa)

SFRM series
21.0 MPa

PFRM series
21.5 MPa15.2 (0%SF)
19.2 (1%SF)
18.6 (2%SF)
17.6 (3%SF)

Steel, fy

388 MPa18.2 (No Fiber)
23.7 (Synthetic)
21.8 (Cellulose)
20.8 (Syn. w/ FA)
16.4 (Cel. w/ FA)
378 MPa

The average maximum recorded axial force(P max ) is used as the basis of comparison of these
retrofitting materials. It was observed that when thestrain reached 0.003, the maximum load was alreadyattained. Tabulated in Table 3 are average maximum
loads (P max ) for each case investigated.

Fig. 5 Comparison of cracks of column with PFRM
jacket (left) and with SFRM jacket (right)

Table 3 Average maximum load of the columns
SeriesCase

SFRM

Control
1% SFRM
2% SFRM
3% SFRM
Control
Plain Mortar
Synthetic
Syn. w/ FA
Cellulose
Cel. w/ FA

PFRM

P max
(kN)
369.5
401.5
437.9
390.4
697.8
673.7
894.7
853.2
794.3
730.2

ANALYSIS OF DATA

% Increasein P max
0.0
8.7
18.5
5.7
0.0
-3.5
28.2
22.3
13.8
4.6

The load that can be carried by control
specimens (columns without jacket) may be
expressed as Eq. (1). Considering the nominal or
maximum loads obtained from tests, the loadcarrying efficacy of concrete (represented by a) may
be calculated. The results are shown in Table 4. It
can be seen that the values obtain for a is very close
to the 0.85 value that is commonly used in design
formulas.

Pmax = a fc' Ac + fy As(1)

Table 4 Efficacy of concrete, a

The increase in P max (expressed in %) of eachretrofitted column is obtained with respect to the
P max of the control specimens of the series. The
largest increase in strength is 28.2%. This is for thecase of synthetic PFRM jacket. For SFRM series,

Type
SFRM
PFRM
452

P max
(kN)
369.5
697.8

α fc’Ac
(kN)
302.4
696.6

fy As
(kN)
121.9
118.8

a
0.82
0.83

GEOMATE- Bangkok, Nov. 14-16, 2016

The calculation results using this model arecompared to the experiment results and are shown in
Fig. 7. It can be seen that relatively good agreement
is obtained between the calculation and theexperiment results. Using this model, the calculated
P max are 369.9 kN and 690.0 kN for the control
specimens of SFRM series and PFRM series,
respectively. These are very close to the test results.

Simulation Using a Simple Model
To simulate the behavior of columns a simple
model is used based on the stress-strain relationshipsof the materials. Shown in Fig. 6 is the stress-strain
curve of concrete and steel. For concrete (shown as
solid line in Fig. 4), the exponential stress-strain
curve developed by Lejano [12] was used as aconstitutive model for concrete. The equation for
the ascending branch of the stress-strain curve of
this model is shown in Eq. (2). The descending
branch is linear from peak stress up to the ultimate
strain of 0.003.

[

fc = fc' 1 − ((ε p − ε ) / ε p )b]Calculation of Confinement EffectAs indicated by test results, there is no significant
increase in axial force because of the enlargement of
the section due to the jacket. In fact, if one tries toinclude the axial force carried by the jacket, the result
would be very much greater than the experimentallyobtained P max . Hence, the increase in axial force may
be concluded to be due to the confinement effect onlyof the jackets. The axial force due to confinement(which is termed as confinement effect) may beobtained by subtracting the calculated strength of thecontrol column specimen from the experimentallyobserved P max . This results to Eq. (4), where C f is theconfinement effect.

(2)

Where fc = stress of concrete at strain (ε), fc’ =compressive strength of concrete, ε p = strain at peakstress = 0.0018 + 0.00001 fc’, and b = 2 -0.0125 fc’.

C f = Pmax − (a fc' Ac + fy As )The confinement effect is evaluated for allcolumn specimens. The results are shown in Table 5.
By dividing C f by the force resisted by the concrete,that is, afc’Ac, then the increase in fc’ may be
determined. This increase in fc’ is denoted as λ andis shown also in Table 5.

Fig. 6 Stress-strain curve of concrete and steel rebar
For steel, the relationship is simply elastic-plastic
model. The steel stress (fs) is linear and proportional
to the strain at the ascending branch, and become
fully plastic after the yield point, that is, fs=fy.
Using these stress-strain relationships, the axial
force (P) at any particular strain of the control
column specimens may be calculated by changing Eq.
(1) into Eq. (3). Note that a is multiplied to fc up to
the peak stress and afterwards the stress ismaintained constant, as shown in the dashed curve inthe stress-strain curve of concrete in Fig. 6.

P = α fc Ac + fs As(4)

Table 5 Confinement Effect and increase
compressive strength of concrete
Case
Control
1% SFRM
2% SFRM
3% SFRM
ControlMortarSynthetic
Syn. w/ FA
Cellulose
Cel. w/ FA(3)

C f (kN)
-0.4
31.6
68.0
20.5
7.8
-16.3
204.8
163.3
104.3
40.2

in

λ = Inc. in fc’ (%)
-0.2
12.8
27.4
8.3
1.4
-2.9
35.8
28.6
18.3
7.0

It can be seen that the FRM provided significant
confinement. Among all cases, the synthetic PFRM
jacket provided the largest confinement effect. For
the SFRM series, the highest increase seems to beobtained in the 2% steel fiber. But when the jacket
is made of plain mortar, negligible confinement
effect was observed. Furthermore, the incorporationof fly ash resulted to less confinement effect.

Fig. 7 Comparison of model with test data of control
column specimens
453

GEOMATE- Bangkok, Nov. 14-16, 2016

Lastly, a model for predicting the load-strain
relationship of the column considering theconfinement effect is proposed by simply increasing
the concrete stress by the factor λ. This is expressedin Eq. (5). The plot of the results of the simulation
using the model as compared to the experimentresults is shown in Fig. 7. It can be seen that
relatively good agreement is obtained. The practicalapplication of this model is in the analysis of RC
structure. Columns retrofitted with FRM jacket may
be simulated using this model.

P = (1 + λ ) 0.85 fc Ac + fs As

expansive soils”, Int. J. of GEOMATE, Vol. 4,
No. 2 (Sl. No. 8), June, 2013, pp. 516-521.
[2] Berthet JF, Ferrier E and Hamelin P,“Compressive behavior of concrete externallyconfined by composite jackets. Part A:
Experimental study”, Construction and Building
Materials, Vol. 19, Issue 3, 2005, pp. 223–232.
[3] Binici B, “An analytical model for stress–strain
behavior of confined concrete”, Engineering
Structures, Volume 27, 2005, pp. 1040–1051.
[4] Sakthivel PB, Ravichandran A and Alagumurthi
N, “Experimental and predictive mechanicalstrength of fiber reinforced cementitious
matrix”, Int. J. of GEOMATE, Vol. 7, No. 1 (Sl.
No. 13), Sept. 2014, pp.993-1002.
[5] Cominoli L, Failla, C and Plizzari GA, “Steeland synthetic fibres for enhancing concrete
toughness and shrinkage behavior”, Proc. Int.
Conf: Sustainable Construction Materials andTechnologies, 2007, pp.231-240.
[6] Sorelli L, Banthia N and Plizzari GA. “Crack
growth resistance of thin mortar layers withhybrid fiber reinforcement”, Special Publication
224, Paper N. 12, ACI, 2004, pp. 161-178.
[7] Abdul Awal ASM, Mohammadhosseini H,
Hossain MZ, “Strength, modulus of elasticity
and shrinkage behavior of concrete containing
waste carpet fiber”, Int. J. of GEOMATE, Vol.
9, No. 1 (Sl. No. 17), Sept., 2015, pp. 14411446.
[8] Valerio DN, Galicia KK, Sia JV and Lejano B,“Retrofitting of RC columns with steel fiber
reinforced mortar”, 15th ASEP International
Convention (15AIC), Manila, Philippines, Sep.
28-29, 2011.
[9] Oropel JC, Panaligan GC and Sy JC, “A study
on the confinement effects of fiber reinforced
mortar (FRM) on RC columns”, Thesis, De La
Salle University, 2013.
[10] Lejano B, “Confinement effects of different
retrofitting materials on the axial strength ofreinforced concrete columns”, 7th HNICEM
Conference, Palawan, Phil., Nov. 13-14, 2014.
[11] Demers M and Neale KW, “Confinement ofreinforced concrete columns with fibrereinforced composite sheets - an experimental
study”, Can J Civil Eng, Vol 26, 1999, pp. 226–
241.
[12] Lejano B, “Modeling the stress-strain
relationship of high strength concrete made of
local
aggregates”,
1st
ASEAN
Civil
Engineering Conference (ACEC), Pattaya,
Thailand, March 12-13, 2009.

(5)

Fig. 7 Comparison of model with test data
CONCLUSION
It had been shown that the FRM jackets wereeffective as retrofitting material for RC columns.
Among all cases investigated, the results indicatethat the largest increase in strength is obtained in thecolumn with synthetic FRM jacket. The axial load
was increased by as much as 28.2%. The
corresponding increase in the confined concrete
strength was evaluated as 35.8%. Both occurred forthe column with synthetic PFRM jacket.
Furthermore, it was shown that the confinedconcrete strength may be determined through thecalculation of the confinement effect by evaluating
the contribution of the different materials. A simple
model was presented that can simulate the behavior
of column as affected by confinement.
ACKNOWLEDGEMENTS
The author acknowledges J.C. Oropel, G.C.
Panaligan, J.C. Sy, D.N. Valerio, K.K. Galicia, andJ.V. Sia for their vital roles in the conduct of theexperimental tests. The author is also grateful to thefaculty and staff of the Civil Engineering Departmentof De La Salle University for their encouragement.
REFERENCES[1] Daoud OM, “Damage assessment andstrengthening of R/C building constructed on454

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DEVELOPMENT OF A NEW SOIL-CEMENT BLOCK USINGPRODUCED WATER FROM OIL FIELDS: A PRELIMNARYINVESTIGATION
1

Khalifa S. Al-Jabri1, A. W. Hago1, Mahad Baawain1 and Gyanendra Sthapit2
Department of Civil and Architectural Engineering, Sultan Qaboos University, Oman; 2Habitech Center,
School of Engineering and Technology, Asian Institute of Technology, Thailand

ABSTRACTOne of the prime aims of the sustainable development strategies in many countries is to provide safe andsecure environment for their inhabitants. This goal can be achieved through the use of environmental friendly
building materials, commonly referred to as green building materials and the effective use of waste materials in
useful applications such as construction industry. In Oman, saving the environment by finding alternative useful
methods for re-using different kind of wastes and seeking ways to reduce the emission of harmful gases, andlooking for affordable low-income housing for people are of great concern to the authorities in order to preserveand sustain the environment for the future generations. In order to fulfill such objective, research into the use
sustainable materials in the construction industry is encouraged. Therefore, in the same context, this study
discusses preliminary results from an ongoing research project between Sultan Qaboos University (SQU) andPetroleum Development Oman (PDO) on the use of produced water from Nimr and Marmoul PDO oil-fields inthe development of a new soil-cement block which possesses good strength, good thermal insulation andconsume lower quantity of cement than the conventional one. Soil and water samples were brought from
Marmoul and Nimr sites. The initial results indicated that both soil and water are suitable for production ofcompressed blocks with good mechanical properties.
Keywords: Soil Compressed blocks, Produced Water, Oil-fields, Strength

to improve the mechanical properties and durabilityof the blocks. These blocks are usually used in nonload bearing construction.
Freshwater is a precious and scarce commodity
in all arid regions such as Oman and other Gulf
countries. This commodity becomes increasingly
more precious in these countries as they witness a
rapid development in their urban infrastructure
including buildings, hospitals, homes, and roads.
Water is needed in road construction where it is usedas mixing water for compaction and for dust control.
It is also needed as a component in concrete
mixtures, primarily for the hydration process of
cementitious materials and for curing. Contractors inarid regions, and especially in remote areas of thedesert, are sometimes faced with the problem offinding water of acceptable quality for theirconstruction work. However, plenty of nonfreshwater is produced in the oil fields during oil
exploration. Water produced with oil forms thelargest amount of waste in the entire oil production
process. An oil field typically produces more than
five times the volume of water as oil during its
economic life. In Oman, the PDO production ofwater from oil-fields was nearly 850,000m³/d in2015. Approximately 9 barrel of water is produced
for each barrel of oil. Therefore, PDO is very keen
to seek for alternative viable uses of oil produced

INTRODUCTION
One of the roles of the sustainable development
strategies in the developing countries is to provide
decent welfare to the community through providing
the basic needs to its people such as clean andpotable water, housing, clean environment, etc. ataffordable and cheap means. Therefore in many
developing countries, it is necessary to seek ways toreduce construction costs, especially for low-income
housing, as well as adopting easy and effective
solutions for their repair and maintenance [1]. Thisobjective can be partially achieved through the useof locally available materials in the construction of
low-income housing. Soil is found in abundant
quantities all over the world. Its use in theconstruction industry is significant in applications
such as cement mortars, concrete, finishing, cement
blocks, and soil compressed blocks and bricks.
Compressed earth block (CEB) system is one of the
traditional construction materials that exists in many
developing countries which have proved to besuitable for a wide range of buildings and whichhave a great potential in the construction of
traditional low-cost dwellings. Adam and Agib [1]reported that soil construction methods are used in
80% of urban buildings and exceeds 90% in rural
areas in Sudan. Often a stabilizer material such as
cement or lime is added in small quantities in order455

GEOMATE- Bangkok, Nov. 14-16, 2016

water (oily water) that can help consuming suchlarge quantity of water. The main objective of thisresearch study is to investigate the potential use ofnon-fresh water (brackish and production water)from Nimr and Marmoul PDO oil- fields in thedevelopment of a new soil-cement block which
possesses good strength, good thermal insulation andconsume lower quantity of cement than theconventional one.

test has to be evaluated for any application of such
soils.
Physical properties of soil
Specific gravity for the two soils was carried out
in accordance with ASTM D854. Two samples
from each soil were selected and tested to determine
their specific gravity. The results show that the twosoils have almost the same specific gravity of 2.69 in
average.
The test results show that Marmoul soil has 29%,
18%, and 11%, liquid limit (LL), plastic limit (PL)and plasticity index (PI), respectively, whereas thesevalues were zero for Nimr soil. The results
indicated clearly that Nimr soil contains more
sand/silt and has no plasticity. For Marmoul soil,
liquid limit LL= 29% and plasticity index PI= 11%.
Both values fall within the limits specified by ARS
680 [2] for production of compressed blocks.
However, since Nimr soil has no plasticity, itssuitability has to be assessed by other tests.
Sieve analysis test was carried out to determine
the distribution (grading) of the soil particles fordifferent sizes. To produce a soil-cement block withgood properties, the soil should contains 15% gravel,
50% soil, 15% silt, and 20% clay at the most. Since
soil contains fine particles i.e. silt and clay,
performing sieve analysis test alone is insufficient toclassify the soil. Therefore, to classify Marmoul andNimr soils, three analyses tests were conducted,
sieve analysis test, wet sieve analysis test, andhydrometer test for soil particles passing 75µm. The
particle distribution curves for the two soils wereplotted together with the limits by ARS 680 [2] inFig. 1. It is clear from the figure that, the two soils
have almost the same distribution; both are within
limits specified by the code ARS 680 [2] forcompressed earth blocks but Nimr soil suits wellwithin those limits.
Also, the results showed that, while the quantityof clay in the two soils are almost the same (2.25%
for Marmoul and 2.02% for Nimr), Nimr soil
contains more silt than Marmoul soil (18%compared to 11%). The quantity of clay in the soil
controls the selection and the amount of thestabilizer. For soils with high clay content, only lime
can be used to stabilize the soil. With the low clay
content as in Marmoul and Nimr soils, cement canbe used as a stabilizer prior to compressing the soil.
However, the quantity of the cement stabilizer canbe determined by testing a number of mixes withvarious stabilizer proportions. Soils with higher clay
contents compress more than those with lower clay
content. In the present case, it is expected that
Marmoul soil will require more volume of soil tomake the same block than of Nimr soil, due to its

MATERIALS AND TESTING METHODS
Soil
Two types of soil were provided from PDO oil
fields: soil from Marmoul oil-field and that provided
from Nimr oil-field.
These soils are notcontaminated soil. Field and laboratory tests were
performed on the two soils to assess their suitability
for block manufacturing. Chemical analyses werecarried out on the soil samples from Nimr andMarmoul sites to determine heavy metals, anions
and organics, whereas tests to determine the physicalproperties of the soil which include specific gravity,
Atterberg limits, shrinkage, and grading.
Produced Water
About 1000L of produced water was broughtfrom each site. For the chemical analysis test of the
produced water measurements included certain
impurities that could affect soil-cement blocks suchas: total alkalinity, sulfate content, chloride content,
total dissolved solids and water hardness. Other
parameters such as pH and conductivity were alsomeasured. The water quality was compared with theinternational specifications for producing concretei.e. ASTM C94 and AASHTO157 as well as the U.S.
Environmental Protection Agency (U.S. EPA).
Soil Stabilizer
Ordinary Portland cement (OPC) was used as astabilizer in the production of Compressed Earth
Blocks (CEB).
RESULTS AND DISCUSSION
Analysis of Soil Samples
Chemical analysis of soil
The results obtained from the chemical analysis
test indicated that the chloride and fluoride contents
in both sites soil samples are similar. The sulphate
content in soils samples from Nimr (1000 mg/kg) isrelatively higher than soil samples from Marmoul(800 mg/kg). In general, the content of heavy metals
in Marmoul soil samples is relatively higher than thesoil samples from Nimr. However, the leachability
456

GEOMATE- Bangkok, Nov. 14-16, 2016higher compressibility.

the water samples from both sites (i.e. less than 5mg/L).
Water quality specifications for producing
concrete mixes according ASTM and AASHTO [3])are shown in Table 1. It can be seen that both water
samples from Nimr and Marmoul meet the limits of
total solids (<50000 mg/L) and sulphate (<3000mg/L). However, the chloride content is higher than
the specified maximum limits (>500 mg/L).
Therefore, further analysis on the production ofbricks has to be conducted.

Percentage Passing (%)
100
90
80
70
60
50

Table 1 Water quality specifications for producing
concrete (adopted from Chini et al. [3]).

4030
NIMIR SOIL
LOWER LIMITUPPER LIMITMARMOL SOIL

20
10
0
0.001
Fig. 10.01

0.1
1Sieve Size (mm)10Chemical Limit
mg/L
Sulphate as SO4
Total Chloride as
ClTotal solids
Alkalis as Na2O
eqv.

100

Particle size distribution for Nimr andMarmoul soils.

The shrinkage test measures the shrinkage of thesoil which contains no stabilizer, and the result canbe used to give a rough guide for estimating therequired amount of cement as a stabilizer. Afterseven days of casting, Nimr sample did not show
any shrinkage in the length. No cracks could be
detected on the surface of the specimen. The
Marmoul specimen showed one hairy crack on its
top surface near the middle of its length. At one end
it separated from the mould. The contraction was
about 0.5mm.

Specification
ASTM C94 AASHTO
M157
3000
3000
500
1000
50000
600

50000
600

Production of Soil-Cement Block
Stulz et al. [4] defined compressed earth blocks(CEBs) as “masonry elements, which are small insize with regular and verified characteristics
obtained by the static or dynamic compression ofearth in a humid state followed by immediatedemoulding”. compressed earth block system is oneof the traditional construction materials that exists inmany developing countries which have proved to besuitable for a wide range of buildings and whichhave a great potential in the construction of
traditional low-cost dwellings.
These blocks are composed of earth materials(soil) with their cohesion is due principally to theclay fraction present in both humid and dry states [5].
The properties and cohesion of the CEBs can beenhanced by the addition of a stabilizer. The qualityof these blocks depends on the type of the raw
materials used i.e. the soil and the stabilizer, and on
the steps and expertise in executing various stages ofmanufacturing i.e. the preparation of materials,
addition and mixing of stabilizers and compaction orcompression up to curing stage [5]. In order toproduce soil-cement blocks which possess good
properties, the most optimum constituent materials
should be selected. The best soil suitable for the
production of soil compressed blocks should beconstituted of 15% gravel, 50% soil, 15% silt, and
20% clay. These properties should be determined
before using the block. Also, the quality of thewater used in producing the blocks should beanalyzed especially if non-freshwater is to be usedor the water to be used is of unknown source.

Analysis Results of Water
Results obtained from the chemical analysis tests
demonstrated
that
thesalinity(electricalconductivity) of the produced water samples from
Nimr is higher than the ones from Marmoul. This isevident from the higher TDS values (8300 mg/L in
Nimr water sample compared to 4780 mg/L in
Marmoul water sample). Moreover, the chloride
content in Nimr water samples (4378 mg/L as Cl-) ishigher compared to Marmoul water samples (2136
mg/L as Cl-). The sulphate (as SO4) and hardness (asCaCO3) contents in Nimr water sample (333 mg/L
and 350 mg/L) are higher compared to Marmoul
water sample (120 mg/L and 140 mg/L). However,total alkalinity (as CaCO3) in Nimr water samples(307 mg/L) is lower compared to Marmoul water
samples (762 mg/L). The content of heavy metals is
relatively low in the water samples from both sites(most are below detection limit except Boron which
was in the range of 4 mg/L). Moreover, oil andgrease content was found below detection limits in
457

GEOMATE- Bangkok, Nov. 14-16, 2016

Generally, typical wet compressive strengths of
CEBs are within 4 MPa [1].
Three soil stabilization methods are usually usedin the production of CEBs. Mechanical stabilization
which involves compacting the soil in order toimprove its resistance to shearing, compressibility,
permeability and porosity. Physical stabilization
which concerns with changing soil texture andproperties as well as the use of different duringmethods such as air or moisture curing, and heat
treatment. Chemical stabilization is achieved byenhancing the physicochemical properties of the soilby the addition of chemicals. The most two widelyused stabilizers in the production of CEBs are
cement and lime. In this research cement will be
used as the main stabilizer although the use of othermaterials such as lime or cement kiln dust (CKD)
can be investigated.
Adam and Agib [1], and Oyelami and Rooy [5]reported many advantages of compressed earth
bricks which are in line with the sustainable
development objectives and strategies. In terms ofmaterials efficiency, CEBs use 30% less watercompared to what is used in other conventionalbuilding material production. They are produced
mainly from soil, sustainable resource, which arerecyclable, produce very little harmful air emissions,have zero or low toxicity, are durable and readily
available [5]. CEBs are energy efficient material;
the production of CEBs requires only 1% compared
to a similar volume of concrete. About 36 MJ (10kW h) of energy is required to produce 1m3 of CEB
in comparison with about 3000 MJ (833 kW h) thatis required to produce the same volume of concrete.
Since the density of CEBs is less than that ofconcrete, they possess less thermal conductivity thanother conventional materials in building construction.
Also, compressed earth blocks have much better fire
resistance and sound proof than other parent
materials. CEB is considered as environmentallyfriendly material since it reduces the consumption ofexcessive volumes of cement through the promotion
and use of environmentally friendly earth materials
that utilizes minimal quantities of cement andconsequently decreases significantly CO2 emission
into the atmosphere. Oylemi and Rooy [5] reported
that manufacturing of a ton of cement generates
around 0.55 tons of CO2 and the burning of carbonfuel produces another 0.4 tons of CO2. Some studies[5]
showed that CEB generates about 22kg
CO2/tonne whereas concrete blocks produce, nearly
143kg CO2/tonne, burnt clay bricks, about 200kg
CO2/tonne and perforated concrete blocks 280 to
375 kg CO2/ tonne.
In this research, two types of CEB
manufacturing machines will be used: a manual anda hydraulic press. These machines are manufacturedby the Habitec Center at the Asian Institute ofTechnology, Thailand. Fig. 2 shows the manual

CEB press whereas Fig. 3 shows a block sample
produced using the manual press.

Fig. 2
press.

The manual soil-cement block making

Fig. 3

Typical soil-cement block.

The manual press produces about 300 bricks per
day while the hydraulic brick press can produce
3,000 bricks per day. Water requirement for the
production is 2,250 liters per day. In tropicalcountries, the top layers of the earth (50cm) are notrecommended to be used as the top soil contains
plants and grasses, which are not good for brickmaking. The deeper soil that are mostly sandy, arefound to be extremely good for the manufacturing ofthese bricks. Before mixing, the soil lumps must be
broken up so that the lumps do not remain. Sand
available in Oman could be normally used without
any further processing (maybe with slight sieving).
Hence, in order to produce good quality bricks, thecontent of these fine particles in the soil should be insuch a proportion that the soil and sand content has
optimum water content and clay particles. The
bricks are produced by stabilizing with cement, andare air cured for 24 hour followed by water curing
for 3 weeks. After 3 weeks the bricks are ready touse for the construction works.
The brick making process is entirely a green
process. It doesn’t need furnace to process the bricks
unlike in the conventional process. The bricks needwater as an ingredient as well as for curing. The
cement component of the bricks which are
predominantly composed of clay and sand is lessthan 15% to attain strengths comparable or better
458

GEOMATE- Bangkok, Nov. 14-16, 2016

than conventional bricks. The manufacture of goodquality, durable compressed stabilized soil-cement
bricks requires the use of soil containing fine graveland sand for the body of the brick, together with silt
or clay to bind the sand particles together. An
appropriate type of stabilizer is added to decrease
the linear expansion that takes place when water isadded to the soil sample. The block is selfinterlocking which requires minimum amount ofmortar for block alignment and fixing leading tosaving in cement used and an improvement in thethermal insulation of the wall by reducing thethermal bridges that are created at the mortar’s line.
To study the possibility of using local soil in the
production of cement-soil block, two locallyavailable soils in large quantities were used. Thesesoils are mountain soil and dune sand. Before using
these soils in the manufacturing of the blocks, theirsuitability was assessed using the particle
determination test which determines the soil
particles content.
Fig. 4 shows the particle
determination test of the mountain soil and dunesand. It can be seen from Fig. 4 that dune sandsample contains one type of soil whereas mountainsoil contains 50% soil and 50% clay. Therefore, thetwo soils cannot be used separately to make goodcompressed earth block, but they can be mixed
together or with other soils to produce blocks thatsatisfy the requirements.

filled with the mixture and pressed using thepressing arm. After removing the blocks from themold, they were then moved to the curing area andleft for 24 hours in the open air under the shade.
After 24 hours, the blocks were moved and stacked
under the sun, covered with plastic covers and left tocure until the day of testing. The compressive
strength test was carried out on the blocks after 28
days of curing as shown in Fig. 5.
The results presented in Fig. 5 show that thecompressive strength of the blocks ranged between
0.87 MPa for Mix-02 and 3 MPa for Mix-04. The
maximum strength was achieved for Mix-04 with 4
proportions of mountain soil and 4 proportions ofdune sand.
Generally these mixtures can be
modified in order to obtain better results in the
futureTable 2

Mixture proportions for the soil-cementblock

Mix no.

CementMix-01
Mix-02Mix-03Mix-04

1
1
1
1

Mountain
soil
8
2
3
4

Dune
sand
0
6
5
4

Water
1.5
0.5
0.75
0.75

Compressive Strength (MPa)

3.5
3
2.5
2
1.5
1
0.5
0
(a) Mountain soil
Fig. 4Mix-01 Mix-02 Mix-03 Mix-04(b) Dune sandFig. 5

Particle determination test

Four soil mixture proportions were preparedusing mountain soil and dune sand as shown in
Table 2. Cement was used as stabilizer and kept
constant in all mixtures. Three water proportions
were used ranging between 0.5 and 1.5. The manualpress machine was used in the production of theblocks. Soil, dune sand and cement were mixed
together. Then water was added to the mixture
providing the necessary moist. The press mold was

Average compressive strength of blocks
after 28 days of curing.

The compressive strength results was obtainedusing field compression testing machine which was
supplied by the manufacturer of the manual andhydraulic press machines as shown in Fig. 6. The
correlation curve between the field testing machinereading and the load cell reading is shown in Fig. 7.
It may be seen from this figure that there is an
excellent correlation between the two parameters
459

GEOMATE- Bangkok, Nov. 14-16, 2016

with R2 value of 0.9987. The compressive strength
values were obtained from Fig. 7 by dividing the
failure load by the net area of the block.

conductivity, whereas, tests to determine thephysical properties of the soil included specificgravity, Atterberg limits, shrinkage, and grading.
The study found that the soil and water from Nimrand Marmoul oil-field sites are suitable for
production of compressed blocks.
Since thepercentage of clay in the two soils is small, cement
can be used as stabilizer for the two soils. The
quantity of the stabilizer has to be determined bystudying different mixes containing various cement
proportions. The study concluded that it is possible
to produce soil-cement blocks with good mechanicalproperties using local soil in Oman.
ACKNOWLEDGEMENTS

Fig. 6

The project’s investigators would like to express
their gratitude to Petroleum Development Oman
(PDO) for providing financial support to the project
under contract number: CR/ENG/CIVL/14/09. Also,
our sincere gratitude is extended to the Asian
Institute of Technology (AIT) for the scientificcooperation in this work.

Field compression testing machine.
160

REFERENCES
y = 1.1134x + 5.3618
R² = 0.9987

Load Cell Reading (kN)

140
120

[1] Adam E.A. and Agib, A.R.A., Compressedstablised earth block manufacture in Sudan,
United Nations Educational, Scientific andCultural Organization (UNESCO), UN, 2001
[2] ARS 680: Compressed Earth Blocks Standards:
A publication by the Center of Development
Industry CDI- CRATerre-EAG,1998, ISBN 2906901-19-9. Contains ARS 680:1996 Code ofPractice for the production of compressed earth
blocks, African Regional Standards.
[3] Chini A.R., Muszynski L.C., Ellis B.S.,
Recycling Process Water in Ready-Mixed
Concrete Operations, 1999, Final Report,
School of Building Construction, University of
Florida, Gainesville, USA.

100
80
60
40
20
0
0

50
100
150
Field Testing Machine Reading (kg/cm2)

Fig. 7
Correlation between
machine and load cell readingfield

testing[4] Stulz R., Mukerji K., ILE S.K., für A.T.
Appropriate Building Materials: a Catalogue ofPotential Solutions, 1993, SKAT, Intermediate
Technology Publications, London.
[5] Oyelami C., and Rooy JLV, A review of the use

CONCLUSION
This report presented the results of apreliminary investigation on the development a new
soil-cement block using produced water from oilfields. The study reported the results of chemicaland physical analyses on produced water and soil
samples from Nimr and Marmoul oil-field sites.
Chemical analyses measurements were total
alkalinity, sulfate content, chloride content, total
dissolved solids and water hardness, pH andof lateritic soils in the construction/developmentof sustainable housing in Africa: A Geological
Perspective, Journal of African Earth Sciences,Vol. 119, 2016, pp. 1-16.

460

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

WASTE PLASTICS AS REINFORCING IN STRUCTURALCOMPOSITE MATERIALS: PROPERTIES AND CHARACTERISTICS
Youssef HALIMI1,2, Hayat BOUCHOUM1, Souad ZYADE2, Mohamed TAHIRI1
1: Geoscience Laboratory, Ain Chock’s Sciences Faculty, Km8, Rte d’El Jadida BP5366,
Maarif, Casablanca , 20100 Morocco
2: Technology High School of Casablanca, Km8, Rte d’El Jadida,
ABSTRACTThe synthetic composite materials have grown spectacularly in the last three decades. Indeed, these lighterand very diverse materials could replace more than 50% of articles and objects made on metals. This is mainly
due to their perfect implementation in various possible and adapted forms and the ratio price/optimized
performance. The work that we present reports the implementation of composite materials based on marble
powder, plastic waste and unsaturated polyester resin with different loads: expanded perlite, Calcium carbonate
and kaolin. These multi-phases materials constitute a system characterized upon its: composition, type ofreinforcement and texture, type and rate of the used fillers and loads.
Keywords: Eco-Materials, Composite Materials, Plastic Waste Recovery, Expanded Perlite.

INTRODUCTION

Domestic plastic waste such as packaging, cans,yogurt containers, drums and jerricans can be used
as reinforcement support if cut to small fibers and beincorporated
incompositematerial.
The
reinforcement of a composite material is expected to
improve the structural or mechanical properties.
The shape of the reinforcement characterizes
different types of monolayer. The reinforcement
may be long fibers (oriented or random), short fibers,
particulate fibers (impregnated or mixed with thepaste) or woven fibers.
Then the rheological tests will be performed on
the obtained material to determine the nature of thesynthesized composite material:- Resistance to bending and breaking force.
- The water absorption.
- The apparent relative density.
- Resistance to deep abrasion for unglazed tiles.
- The chemical resistance.

Many solid wastes are abandoned in nature.
Industrial Plastic wastes are in general somehow
dangerous and have long life time (200 -1000 years).
Their degradability needs a long-time.
Giving a second life to used plastic is the aim of thisstudy. In this work, we focus on the valorization of
marble powder and recycling of plastic waste (bags,
bottles, jars, cans ...), while having a purely
environmental (recycling and recovery), economic(without energy) and ecological character (ownprocess).
The idea is to use plastic wastes (Synthetic sand) as
reinforcement for composite materials based on
marble powder, and unsaturated polyester resin withdifferent fillers: expanded perlite, talc, calcium
carbonate, and kaolin.
MATERIALS AND METHODS
It is to synthesize a polymer material based on
unsaturated polyester resin by polymerization withthe catalyst called PMEK (Peroxide Methyl Ethyl
Ketone) and a hardener like Cobalt octoate.
The polymer is filled with inorganic materials,
for example, marble powder, expanded perlite,
calcium carbonate, kaolin, synthetic sand, talc, etc.

RESULTS AND DISCUSSION1-FAMILY COMPOSITION:

Model of sample’s composite materials synthesized:

461

GEOMATE- Bangkok, Nov. 14-16, 2016••

Family 1
MatrixReinforcement

Insaturated polyster
resin
Bags of plastic(Impregnatedparticulate
reinforcement)

Family 2’’’25 %MatrixUnsaturated polyester
resin

30 %

25 %

Reinforcement

Bags of plastic
Oriented fibrous
reinforcement UD

10 %ChargeMarble powder

46 %ChargeMarble powder

56 %Accelerator

Octoate of cobalt

3%

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

Catalyst

Methyl ethyl ketone
peroxide1%••

Family 2

Family 3

MatrixInsaturated polyester
resin

25 %

MatrixUnsaturated polyster
resin

25 %

Reinforcement

Bags of plastics
Oriented fibrous
reinforcement UD

25 %Reinforcement

Bags of plastics
Woven reinforcement

25 %ChargeMarble powder

46 %ChargeMarble powder

46 %Accelerator

Octoate of cobalt

3%

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

Catalyst

Methyl ethyl ketone
peroxide1%••

Family 2’
Unsaturated polyester
resin
Bags of plastic
Oriented fibrous
reinforcement UD

Family 4

ChargeMarble powder

51 %ChargeUnsaturated polyester
resin
Bags of plastic
Random fibrous
reinforcement
Marble powder

Accelerator

Octoate of cobalt

3%

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

Catalyst

Methyl ethyl ketone
peroxide1%

MatrixReinforcement

•

25 %

Matrix

20 %Reinforcement•

Family 2’’

25 %
25 %
46 %

Family 5

MatrixUnsaturated polyester
resin

27 %

MatrixReinforcement

Bags of plastic
Oriented fibrous
reinforcement UD

15 %Reinforcement

ChargeMarble powder

54 %ChargeMarble powder

46 %Accelerator

Octoate of cobalt

3%

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

Catalyst

Methyl ethyl ketone
peroxide1%

462

Unsaturated polyester
resin
Bags of platic(particulate
reinforcement mixed
with the paste)

25 %

25 %

GEOMATE- Bangkok, Nov. 14-16, 2016••

Family 6
MatrixUnsaturated polyester
resin

26 %

Reinforcement

Glass woven fabrics-

ChargeMarble powder

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%•

Family 10
MatrixReinforcement
ChargeAccelerator
Catalyst•

Family 7
MatrixUnsaturated polyester
resin

26 %

Reinforcement

Canvas woven fabrics-

ChargeMarble powder
Octoate of cobalt

Catalyst

Methyl ethyl ketone
peroxide

MatrixReinforcement•

3%
Accelerator
1%

Catalyst•

Family 8
Unsaturated polyester
resin

26 %

MatrixReinforcement

Glass woven mat-

Reinforcement

ChargeMarble powder

70 %ChargeOctoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%•

Accelerator
Catalyst•

Family 9
Unsaturated polyester
resin

21 %

MatrixReinforcement

Bags of lastic
Woven reinforcement

25 %Reinforcement

ChargePerlite50 %Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide463

Marble powder +Kaolin36 % +
10 %Octoate of cobalt
Methyl ethyl ketone
peroxide3%
1%

Unsaturated polyester
resin
Plastic pots, bottles,
cans Cut. UD oriented
fiber reinforcement
Marble powder +PerliteOctoate of cobalt
Methyl ethyl ketone
peroxide26 %25 %30 % +
15 %3%
1%

Unsaturated polyester
resin
Plastic pots, bottles,
cans Cut. UD oriented
fiber reinforcement
Marble powder +synthetic sandOctoate of cobalt
Methyl ethyl ketone
peroxide24 %25 %37 % +
10 %3%
1%

Family 13

Matrix1%

25 %Family 12

MatrixAccelerator

25 %Family 11

ChargeAccelerator

Unsaturated polyester
resin
Plastic pots, bottles,
cans Cut. UD oriented
fiber reinforcement

Unsaturated polyester
resin
Plastic pots, bottles,
cans Cut. UD oriented
fiber reinforcement

25 %
25 %

ChargeMarble powder +Calcium Carbonate

36 % +
10 %Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

GEOMATE- Bangkok, Nov. 14-16, 2016•

MatrixUnsaturated polyester
resin

25 %

Reinforcement

Plastic pots, bottles,
cans Cut: Woven
reinforcement

25 %ChargeMarble powder +KaolinAccelerator

Octoate of cobalt

Catalyst

Methyl ethyl ketone
peroxide•

The flexural strength increases with increasing thefiber percent. This is due mainly to the adhesion
matrix - fiber and the ability to stop the spread ofcracks by fibers.

Family 14

Table 2: Effect of reinforcement architectureN°sample

31 %
+
15 %
3%

1

1%
2

Family 15
MatrixUnsaturated polyester
resin

27 %

3

Reinforcement

Plastic pots, bottles,
cans Cut: Woven
reinforcement

25 %

4

ChargeMarble powder

44 %

5

Accelerator

Octoate of cobalt

3%

Catalyst

Methyl ethyl ketone
peroxide1%

Reinforcement
architecture
Particulate
reinforcement (Veryfineparticles):ImpregnatedFiber reinforcement(Long fibers):
Oriented(Unidirectional)
Mat fabricreinforcement
Fibrousreinforcement (longfibers) : Random
Particulate
reinforcement (veryfine particles) :mixed with paste

Tensilestrength(N)

Stress(MPa)2310.5

29.75

2380

30.64

2520

32.44

2490.931.93

2360.4

30.39

DETERMINATION OF FLEXURAL STRENGTH
2AND TENSILE STRENGTH:

Table 1: Effect of reinforcement rate
N°
Sample
2
2’
2’’
2’’’

Reinforcement
rate
25 %
20 %
15 %
10 %

Tensilestrength (N)
2380
2345
2290.12275.6

Stress
(MPa)
30.64
30.19
29.5
29.3

Fig 2. Determination of flexural strength of samples(1, 2, 3, 4, 5)
The mechanical properties of the composites
materials are influenced by the fiber architecture.
The fabric texture always has the best strength:
Fabrics have surface properties that give them
resistance in two dimensions. However the rate ofincrease of this property is highly dependent on thearchitecture of fibers introduced in matrix. Thus themechanical properties of composites depend
strongly on the direction.
If the reinforcements are randomly distributed, theobtained material has average performance; while if
the fibers are oriented in a specific direction, theproduced material has a resistance which depends onthe direction of the applied load.

Fig 1. Determination of flexural strength of samples(2, 2’, 2”, 2”’)464

GEOMATE- Bangkok, Nov. 14-16, 2016

The bending strength in the fiber direction is greaterthan that perpendicular to the fibers. So it isnecessary to provide layers of crossed fibers toensure interesting properties in different directions.
Also, this allows adapting the material to theconsidered piece by making more fiber in the
direction of the principal forces to which the part issubjected.
Table 3: Reinforcement of simplesN°
sample3678Type ofreinforcement
Plastic bags as
tissueGlass fibers
TissueMat of glass
fibersTensilestrength(N)
2520

Stress(MPa)2623.1
2500.23
2583.12

33.77
32.19
33.26

Fig 5. Determination of flexural strength of samples(10, 11, 12, 13, 14)
The addition of an inorganic phase (kaolin, calcium
carbonate, perlite, synthetic sand) to the polymer,
leads generally to an increase in stress. The material
properties are a function of those of each component,
but also the proportions in which they operate.

32.44

Table 5: Effect of thickness
Thickness
(mm)
10.4
20.3

Tensile strength(N)
2380> 6137.5

Stress
(MPa)
30.64
-

The bending strength increases with increasingmaterial thickness.
Table 6: Determination of water absorption
Reinforcement architecture
Particulate reinforcement (Very fineparticles) : ImpregnatedFibrous reinforcement (long fibers):
Oriented (Unidirectional)

Fig 3. Determination of flexural strength of samples(3, 6, 7, 8)
The used process allows obtaining reinforcementwith plastic bags of tissue similar to that obtained by
the canvas fabrics. Also, the properties arecomparable to other technical fabrics: glass fabricand glass mat.

Type of charges
Marble powder+
KaolinMarble powder+
PerliteMarble powder+
Synthetic sandMarble powder+
Calcium
Carbonate
Marble powder+
Kaolin

36 %
+
10 %
30%
+
15 %
37 %
+
10 %

Tensilestrength(N)

Stress(MPa)2605.733.55

2420.631.16

2552.0932.86

2615.5

33.67

31 %
+
15 %

2755.6

35.5

1.4
1.9

Fibrous reinforcement (Long fibers) :RandomParticulate reinforcement (Very fineparticles) : Mixed with paste

1.8
1.2

Table 7: Determination of apparent relative density:
Reinforcement architecture
Particulate reinforcement (Very fineparticles) : ImpregnatedFibrous reinforcement (long fibers):
Oriented (Unidirectional)
Mat woven reinforcement

36 %
+
10 %

1.0

Mat woven reinforcement

Table 4: Effect of charge
Rate ofcharges

Absorption%

Fibrous reinforcement (Long fibers) :RandomParticulate reinforcement (Very fineparticles) : Mixed with paste

465

Density
2
2.1
2.05
2.03
1.78

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 9: Resistance to deep abrasion of materials:

Composite materials are being increasingly used forreasons of weight gain. These materials allow
achieve lower weight structures to those of metallic
materials, while keeping rigidity and resistancegreater or equal than metallic materials.

Type of charges
Marble powder+
KaolinMarble powder+
PerliteMarble powder+
Synthetic sandMarble powder+
Calcium carbonate

Table 8: Determination of resistance to deep
abrasion:
Reinforcement architecture
Particulate reinforcement (Very fineparticles) : ImpregnatedFibrous reinforcement (long fibers):
Oriented (Unidirectional)
Mat woven reinforcement
Fibrous reinforcement (Long fibers) :RandomParticulate reinforcement (Very fineparticles) : Mixed with paste

3

V(cm )
262
262
227
227

Rate ofcharges
36 %
+
10 %
30%
+
15 %
37 %
+
10 %
36 %
+
10 %

V(cm3)
215

262

262

215

Resistance to deep abrasion of materials is
influenced by the fiber architecture and the nature of
charges introduced into the matrix, kaolin andcalcium carbonate increases the strength comparedto the other compositions.

215

REFERENCES
CONCLUSION[1]

L. Berreur, B. Maillard, S. Nösperger. L’industriefrançaise des matériaux composites : des enjeux
prioritaires pour un développement durable. Etude
Digitip. 2001, 129 pages.
(b) Matériaux composites: Comportement mécanique etanalyse des structures 3° Ed.
[2] G. Chrétien. Matériaux composites à matrice organiques.
Lavoisier, 1986, Partie 1, pp11-52.
[3] G. Chrétien. Matériaux composites à matrice organiques.
Lavoisier, 1986, Partie 2, 53-106.
(b) Pratique des Plastiques et Composites, Ouvrage
Collectif, Référentiel Dunod. 1999.
[4] G. Chrétien. Matériaux composites à matrice organiques.
Lavoisier, 1986, pp495(b) Durabilité des matériaux composites à matriceorganique, Revue des composites et des matériaux
avancés. 2002, 12, N° 1. (c) J-A. Manson, L. Carlsson, JP. Mercier, P-E. Bourban. PPUR. 2004, 18, 234 pages.
[5] N. Bahlouli. Composites. Cours DESS MASI, Universitélouis Pasteur de Strasbourg, 2002
[6] M.Nardin, Interfaces fibre-matrice dans les matériaux
composites. Applications aux fibers végétales, Revue desComposites et des Matériaux Avancés. 2006, 16, 1.
[7] (a) M. Reyne. Technologies des composites, Editions
Hermes.1995.
(b) P. Keflouan, F. Boterel, J. Lostec, J-P. Bertotm J-M.
Haussonne. J. Phys. 1991, 1, 1723-1750.
[8] C. Bathias. Matériaux composites. Dunod, Collection
Technique et Ingénierie. 2005, 432 pages.
[9] Fiche technique de perlite, PERLITE INC. Imm. 4
Saison, Colline 20190 Casablanca[10] JEAN BOST : Matières plastiques I chimie- applications.
Edition TEC et DOC[11] MAURICE REYNE : Les composites. Edition : que saisje 1995. pp: 18 ; 19- 21[12] M.Benhaddou, A Haddout, G. Villotreix, ‘’Rheologicalof Polymer Composites PP/FG at high shear rate’’,The composite materials with organic matrix,
reinforcement in short filaments of used plastic bags
(waste) or cut plastic bottles and mineral charges(expanded perlite, marble powder, ...) are a family ofenvironmental friendly materials and the way ofsustainable development projects. In fact, thesekinds of eco-materials can have various applications
in buildings as walls and grounds coating andcovering. The physical, mechanical, and chemical
properties of these eco-materials compared to theircounter family marketed, prove that they can very
well meet the customer requirements and conform tostandard norms.
Moreover, it should be noted that these materials
concern three fundamental aspects:• Economic: Low cost of energy gains forprocessing, the use of recycled products(plastics and worn local marble powder)and the ease of the manufacturing process.
• Technical: Mechanical properties
comparable to traditional reinforcements
based materials (glass fiber), significant
reduction in production cycle time, nodeformation and withdrawals, low density
and chemical stability for different reagents.
• Ecological: allows the reuse and recyclingof waste, requires no energy to implement,
can be installed throughout the territory inan area near the landfill if necessary

466

GEOMATE- Bangkok, Nov. 14-16, 2016

Édition : Service de normalisation industrielle marocaine
2000.
[16] Norme Marocaine MN ISO 10545-6 : Détermination dela résistance à l'abrasion profonde pour les carreaux nonémaillés. Édition : Service de normalisation industrielle
marocaine 1995.
[17] Norme Marocaine NM ISO 10545-13 : Détermination dela résistance chimique. Édition : Service denormalisation industrielle marocaine 2000.

International Journal of Reinforced Plastics andComposites , Vol.26, N°13, 2007[13] Caractérisation des matières premières, Laboratoire
public d’essais et d’études (LPEE) – Maroc 2002
[14] Norme Marocaine MN ISO 10545-4 : Détermination dela résistance à la flexion et de la force de rupture.
Édition : Service de normalisation industrielle marocaine
2000.
[15] Norme Marocaine MN ISO 10545-3 : Détermination del’absorption d’eau, de la porosité ouverte, de la densité
relative apparente et de la masse volumique globale.

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

MODELLING OF CARBONATION OF REINFORCED CONCRETESTRUCTURES IN INTRAMUROS, MANILAUSING ARTIFICIAL NEURAL NETWORKRichard M. De Jesus1, Joshua A. M. Collado2, Jemison L. Go3, Mike A. Rosanto4, and John L. Tan5
1
Faculty of Civil Engineering in De La Salle University (Manila), Philippines
2-5
Undergraduate Student of Civil Engineering in De La Salle University (Manila), Philippines

ABSTRACT
Corrosion is a perennial problem in reinforced concrete structures, and is a serious concern due to thedeterioration that it causes to reinforced concrete members. Though regarded as having a minor influence tocorrosion compared to chloride-induced corrosion, carbonation is becoming a serious threat due to continuous
development of cities like Manila. Expectedly, as Manila continues to develop, carbon emission shoots up toalarming proportions, calling out for studies to investigate and mitigate its effect to human health and structures.
Artificial Neural Network (ANN) is known for establishing relationships among parameters with unknown
dependency towards another variable, similar to the case of carbonation’s dependency with age, temperature,
relative humidity, and moisture content. Utilizing field-gathered secondary data as training and testing parameterfor back propagation algorithm, an ANN model is proposed. Prediction of carbonation depth using ANN Model
C421 showed reliable results. Validation of performance of Model C421 was further checked by comparing its
prediction with a different set of field-gathered secondary data and results confirmed good agreement between
prediction and measured values.
Keywords: Carbonation, Artificial neural network (ANN), Model C421, Reinforced concrete

study, the level of CO2 has reached 90.78 million
metric tons in 2013. By the beginning of 2015, CO2
emissions has reached 97.91 million metric tons [5].
With this alarming rate of increasing CO2 levels
in Metro Manila, it is not only human health that is
at risk, but the health of reinforced concrete (RC)structures as well. CO2 penetration in RC structures,
known as carbonation, poses a great threat to thestructural health as its risk to corrode reinforcing
steel bars (rebar) is high. Structures in urban/cities
are most vulnerable to carbonation due to high levelsof CO2. And to raise the concern farther, only
limited studies were done in the country thatinvestigates the effect of CO2 in RC structures, andhow fast it contributes to the deterioration of
structures. This deterioration happens within RC
members are hard to detect from the outside whichmakes it threat even more alarming.

INTRODUCTION
Recent years had shown Philippines’ economy
rising fast. As of first quarter of 2016, Philippine
economy grew close to 7% and considered to be thehighest among 11 Asian countries as reported in [1].
The same news agency, on May 16 2016, reported
that Philippines is the top performing country inSoutheast Asia based on the report of Oxford
Business Group (OBG) [1].
According to the Philippine Statistics Authority(PSA), Philippines’ gross domestic product (GDP),for the 1st quarter of 2016, was 6.9 percent. It is
recorded to be higher than that of China’s 6.7% andVietnam’s 5.7% [2].
The country’s economic rise is warmly
welcomed by Filipinos. But it calls for the country toalso grow accordingly with such economic growth.
An indicator that the country is lagging behind withthis growth, at some areas, is the failure to anticipate
overwhelming increase in motor vehicles and its
effects, in the past years. As reported by [3],passenger car sale grew from a little over 20,000 in2005 to over 63,000 in 2015 while that ofcommercial vehicles grew from over 10,000 to over
30,000 in the same period.
A quantification of effects of increasing car
volume, in terms of carbon dioxide (CO2) emission,
was presented in the study of [4]. According to thisTHEORETICAL BACKGROUND
Carbonation in Reinforced Concrete
Carbonation is a chemical process that involves
the reaction of some chemicals in concrete,
specifically the cement hydration elements calcium
hydroxide (Ca(OH)2), and calcium-silicate-hydrates(C-S-H), with CO2 from the atmosphere thatpenetrates into concrete. The chemical reaction
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GEOMATE- Bangkok, Nov. 14-16, 2016

produces calcium carbonate (CaCO3) which canpotentially lower the passivity of steel bars, hence,
contributing to corrosion. This chemical process canbe represented by the following chemical equation:
CO2 + Ca(OH)2  CaCO3 + H2O

claimed that thin layer of water covering the pores
tend to block the intrusion of CO2.

Artificial Neural Network(1)
Artificial Neural Network (ANN) is an effective
tool for prediction of behavior or phenomenon.
Unlike other tools, it requires no assumption as to
how independent variables are affected by a single
or multiple independent variables. The only basic
requirement it requires is sufficient amount of data.
ANN can be seen as a black box where input data
are being fed to produce an output data [12]. ANN isefficient and effective in modelling systems
influenced by multiple variables with unestablishedinterrelationships especially with incomplete data
[12]. Reference [13] described ANN as made up oflarge number of highly interconnected processing
elements called neurons which work in unison tosolve a problem.
Before ANN can establish multi-variableinterrelationships and come up with a prediction, it
needs to learn the system first which can be achievedby feeding input data with set initial conditions.
Reference [12] added that ANN requires nofunctional relationship among variables since theprocess is data driven implying that it can adopt tothe training data and captures the relationshipbetween the input and output variables.

The study of [6] stated that carbonation involves
the dissolution of CO2 with water in the pores ofconcrete, before reacting with Ca(OH)2, to form
carbonic acid (CaCO3). A quantification of how
much of CO2 that penetrates into concrete is
attributed to carbonation is explained by [7]. It
claimed that only half (50%) of CO2 reacts with
Ca(OH)2. The other half reacts with C-S-H.
The rate of how fast carbonation progresses in
concrete is not constant [8]-[9]. This is attributed tocontinuous hydration of cement particles in concrete
and with continuous reaction of CO2 with Ca(OH)2.
With less Ca(OH)2 available, less reaction with CO2
occurs. The continuous hydration also densifies the
microstructure of concrete which lowers the
carbonation diffusion rate further.
Several factors were identified to influencecarbonation. Major factors were water to cement
ratio, degree of hydration, CO2 concentration of thesurrounding air, moisture content, temperature,alkali content, and presence of damaged zones andcracks [10].
This study, however, will only consider age,temperature, relative humidity, and moisture content
in modelling carbonation through artificial neuralnetwork (ANN) due to availability of data.
Obviously, age is of primary importance in termsof influence to carbonation. With lengthy exposure
to CO2, and extended duration of diffusion of CO2
in concrete, carbonation depth increases. It is
assumed in this study that Fick’s law of diffusion
which relates carbonation depth with the product ofcarbonation coefficient and square root of time, isvalid.
As temperature increases, the diffusivity ofgaseous CO2 increases [8]. The diffusivity is seen as
the rate of CO2 to react and fully blend with cement
hydrating elements. The study further added that therise in diffusivity can be attributed to increased
molecular activity. According to [11], the rise intemperature translates to higher solubility ofchemical compounds. This is seen as a process that
potentially facilitates the carbonation process faster.
Relative humidity is seen to contribute tocarbonation in an opposite manner compared to thatof temperature. Higher relative humidity lowers therate of carbonation according to [8]-[9] due topresence of high moisture in concrete pores. Since
diffusivity of CO2 is lower in water than in air, it isanticipated that CO2 diffusion in concrete pores
lowers with increasing humidity. Moreover, [8]-[9]Neural Network Modelling, Data, and Results
Data and its preparation for ANN modelling
A secondary data, based from [14] was utilized
for this study for the training of the neural networkmodel. The validation of model was conducted using
another secondary data based on the study of [4].
Both studies conducted carbonation test to structures
within the same vicinity – Intramuros, Manila.
Normalization of data was performed for these
two secondary data. Normalization is to address thewide gap in terms of magnitude between fieldmeasured carbonation data and that of the input
variables. The process was done by simply getting
the quotient between the data and the maximum
value for that particular parameter. This process was
implemented to all parameters – age, temperature,
relative humidity, moisture content, and carbonation
depth. After normalization, all data range from 0 to 1.
Choosing the ANN architecture
A sample ANN architecture is shown in Figure 1.
The first 2 brown circles (a.k.a. neurons) are theinputs and could vary from 1 to N depending on thenumber of variables considered to influence an
output (represented by red circle). The two blue469

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1. Comparison of six ANN architecture

circle in between represents the neurons in thehidden layer.
The arrows connecting the neurons havecorresponding weights which change as ANN“learns” the relationship between input and output.
This sample ANN architecture represents a feedforward back propagation network which was
adopted in this study.

ANN
Model

error

training R

testing R

C411
C421

0.034

0.811

0.967

0.011

0.942

0.969

C431

0.010

0.949

0.922

C441

0.015

0.912

0.975

C451

0.014

0.923

0.989

C4101

0.016

0.931

0.959

The measured carbonation depths as presented in Fig.
2 to 7 were taken from field-measured carbonation
depth.

Figure 1. A sample neural network
The neural network architecture adopted in thisstudy considered 4 inputs – age, temperature,
relative humidity, and moisture content; while
output was set at 1 – carbonation depth. One hidden
layer was also considered in this study as more
hidden layers tend to make the training and learningwith ANN more difficult and complex. The numberof neurons in the hidden layer, however, was varied
from 1 to 10 based on the recommendation of [15].
Choosing the number of neurons, for the final ANN
model, in the hidden layer was based on thefollowing criteria: (1) the training R must be 0.80 or
higher; (2) the testing R must be 0.90 or higher; and(3) error must be lower than 5% (or 0.05);
Tansig (or log sigmoid) transfer (or activation)function was adopted in the study with LevenbergMarquadt (LM) as the training algorithm. The rest of
the training parameters used were the default values
in MATLAB – the software used to execute ANN
process.
Comparison of these 3 criteria for 6 considered
ANN architecture was summarized in Table 1. Six
ANN architectures, of varying number of neurons in
the hidden layer (n =1, 2, 3, 4, 5, & 10), werecompared based on the above criteria. Based on all 3
criteria, all architectures met the requirements.
Ranking of ANN architectures based on the 3
criteria showed that Model C421 was always in the
top three (3). Thus, it was decided that Model C421
will represent the modelling of the carbonation depth
using ANN.
Prediction of carbonation depth against measured
values is shown on Fig. 2 to 7. These show slightdeviation between prediction and measured values
showing that ANN closely established therelationship between carbonation depth and thecarbonation parameters (age, temperature, relative
humidity, and moisture content).

Figure 2. C411 prediction v. actual

Figure. 3 C421 prediction v. actualComparison between measured carbonation
depths and the prediction of Model C421, as shownin Fig. 3 showed minimal variation. ANN model
C421 is considered to have learned the carbonation
behavior via Artificial Neural Network (ANN). How
carbonation depth varies with age, temperature,
relative humidity, and moisture content is closely
predicted by C421 model.

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GEOMATE- Bangkok, Nov. 14-16, 2016

Figure 4. C431 prediction v. actual

Figure 7. C4101 prediction v. actual
Validation of Model C421
To further evaluate the performance of Model
C421, it is tested against the carbonation data
gathered from the study of [4]. Raw data from thisstudy were also normalized, similar to the processadopted for [14]. ANN model C421 were adopted
and its optimized weights (shown in Table 3) wereset as the initial weights for training the network
using MATLAB. Its biases, shown in Table 2, werealso adopted.

Figure 5. C441 prediction v. actual
Since all models showed close values tomeasured ones, choosing C421 on the basis of
testing R criteria is supported.

During validation, C421’s training R is 0.9831
while its testing R is 1.0. The results of validation
can be seen in Fig. 8. This result confirmed that
Model C421 approximates the carbonation
prediction with high reliability.

Table 2. Initial Biases for C421 Model
Input to hidden-0.003-3.012

Hidden to ouput
0.395

Table 3. Initial Weights for C421 Validation
nodes
Age
Temp
R.H.
m.c.
Depth
1
0.6
0.72
0.078 -1.07
3.30
2
-0.2
-8.97
-1.61
5.00
3.25
Note: nodes refer to hidden nodes, R.H is relative
humidity, m.c. is moisture content, and depth is the
carbonation depth
Figure 8. C421 validation
CONCLUSION
The country’s remarkable economic growth has
been applauded locally and internationally. But
some of the environmental effects of a rising
economy are treated with less attention. Amongthese is the alarming increase in CO2 levels in theenvironment and the potential hazard it poses toreinforced concrete structures due to carbonation.
There are few studies conducted locally toinvestigate, and analyze the effect of CO2 intrusion

Figure 6. C451 prediction v. actual471

GEOMATE- Bangkok, Nov. 14-16, 2016

in RC structures. This study was among the initialsteps to understand the carbonation phenomenon.
Here, CO2 intrusion in concrete, throughcarbonation depth, is related to age, temperature,
relative humidity, and moisture content using feedforward backpropagation artificial neural network.
ANN model C421 were trained using asecondary data and found close prediction values
with measured carbonation depth. This was further
validated by comparing the prediction values withmeasured values from another secondary data.
Results showed that model C421 were able tovalidate the reliability and accuracy of the model as
depicted by close values between prediction andmeasured carbonation depths.

[6] Burkan Isgor, O., and Razanqapur, A. G. (2004).
Finite element modeling of couple heat transfer,moisture transport and carbonation processes in
concrete structures. Cement and Concrete
Composites, 26(1), 57-73.
[7] Park, D. C. (2008). Carbonation of concrete inrelation to CO2 permeability and degradation of
coatings. Construction and Building Materials,
22(11), 2260-2268.
[8] Talukdar, S., Banthia, N., & Grace, J. R. (2012).
Carbonation in concrete infrastructure in thecontext of global climate change – Part 1:
Experimental results and model development.
Cement and Concrete Composites, 34(8), 924930.
[9] Yoon, I. S., Copuroglu, O., & Park, K. B.
(2007). Effect of global climatic change on
carbonation progress of concrete. Atmospheric
Environment, 41(34), 7274-728.
[10] Houst, Y. F. & Wittman, F. H. (2002). Depth
profiles of carbonates formed during naturalcarbonation. Cement and Concrete Research, 32(12), 1923-1930.

ACKNOWLEDGEMENTS
The author would like to thank Engr. Ronaldo S.
Gallardo for allowing the use of data from theirstudy in the development (& training) of the ANN
model in this paper.
REFERENCES[1] Jiao, C. (2016, May 16). PH is best economy inSoutheast Asia – Oxford Business Group. CNN
Philippines. Retrieved on June 11, 2016 fromhttp://cnnphilippines.com/business/2016/05/13/
oxford-business-group-economy-philippinessoutheast-asia.html[2] Remitio, R. (2016, May 19). PH is fastestgrowing economy in Asia, expands by 6.9% in
Q1. CNN Philippines. Retrieved on June 11,
2016
from
http://cnnphilippines.com/news/2016/05/19/philippines-fastest-growing-economy-asia-gdpq1.html[3] Punongbayan, J. C., & Mandrilla, K. (2015,
August 15). Carmaggeddon: Why are there somany cars in Metro Manila?. Rappler. Retrievedon June 11, 2016 from http://www.rappler.com/
views/imho/102701-carmageddon-metro-manila[4] Collado, J., Go, J.L., Rosanto, M.A., Tan, J.L.,and De Jesus, R.M. (2015). Employing
Artificial Neural Network in Assessing Various
Factors Affecting Carbonation in old structures.
Thesis. De La Salle University, Manila.
[5] Philippine Carbon Dioxide Emissions. (2013).
Retrieved on June 11, 2016 from https://ycharts.com/indicators/philippines_carbon_diox
ide_emissions.

[11] Kolio, A., Pakkala, T. A., Lahdensivu, J., &Kiviste, M. (2014). Durability demands related
to carbonation induced corrosion for Finnish
concrete buildings in changing climate.
Engineering Structures, 62-63, 42-52.
[12] Oreta, A. W. C. (2004). Simulating size effecton shear strength of RC beams without stirrups
using neural networks. Engineering Structures,
26(5), 681-691.
[13] Lu, C., & Liu, R. (2009). Predicting carbonation
depth of prestressed concrete under different
stress states using artificial neural network.
Advances in Artificial Neural Systems,2009,1-8.
[14] Bata, K. J., Beltran, B., Sacaguing, A., &Gallardo, R. (2013). Carbonation profile of oldand new reinforced concrete structures inIntramuros, Manila. Thesis. De La Salle
University, Manila.
[15] Hect-Nielson, R. (1989). Theory of backpropagation neural networks. Proceedings of theInternational Joint Conference on Neural
Networks, 593-611.

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INFLUENCE OF 3D NUMERICAL MODEL PARAMETERS IN
ANALYSES OF THE SUBSOIL-STRUCTURE INTERACTION
1,2

Jana Labudkova 2, Radim Cajka 1
Faculty of Civil Engineering, VSB - Technical University Ostrava, Department of Structures, Ludvika
Podeste 1875/17, 708 33 Ostrava – Poruba, Czech Republic

ABSTRACT
For decades attention has been paid to interaction of foundation structures and subsoil and, in turn, todevelopment of interaction models. Complexity of a static solution is given mainly by selection of acomputational model, effects of physical-nonlinear behaviour of such structure and co-effects of the upper
structure and the foundation structure. ). Input data for numerical analysis were observed experimental loading
test of steel-fibre reinforced concrete slab. The loading was performed using unique experimental equipment
which was constructed in the area Faculty of Civil Engineering, VŠB-TU Ostrava. Homogeneous half-space thistakes no account and calculated settlement is strongly dependent on the size of the subsoil model, as parametricstudy demonstrated. The modulus of deformability changes continuously, depending on the depth, in theinhomogeneous half-space. Values calculated by 3D numerical model were compared with values measured
during the loading test of steel-fibre concrete slab.
Keywords: Foundation structure, Soil – structure interaction, Contact stress, 3D FEM element.

in the steel-fibre reinforced foundation slab were
evaluated. Another experimental loading tests andtheir results are also described in [2, 3, 6, 7, 8, 11, 19,
21, 22].

INTRODUCTION
Input characteristics of applied material models
in subsoil-structure interaction often have lower
levels of reliability (especially in the case of the
subsoil). This can lead to a lower level of reliabilityof results of modeling. For this purpose, researchand experimental measurements focused on thesettlement of loaded foundation soil, deformations of
the slabs in interaction with subsoil and dependenceof stress in the slabs on the characteristics of the
subsoil are still performed. Because calculated
subsidence and real subsidence of foundations do
not correlate well [9, 10, 17, 18, 20], a site survey isneeded and experimental measurements are carriedout in order to determine subsidence of foundationsoil under structures, deformation of foundation
slabs and characteristics of stress in foundation slabs
which depend on parameters of subsoil. Using
results of such experiments, the methods used for
calculation of subsidence are modified and become
stricter. In 2010, testing equipment – a stand, waserected in the Faculty of Civil Engineering, VŠB –TU Ostrava [15]. The stand measures deformationand monitors interaction between stress anddeformation. In 2014 an experiment was carried out
using the stand. Values measured during the loading
tests were compared with values calculated byinteraction 3D FEM models. The calculations werecarried out for several sizes of the subsoil and for
different boundary conditions. The values werecompared and impacts of 3D numerical model
parameters on final deformation and internal forces

EXPERIMENTAL LOADINGSTEEL-FIBRE REINFORCEDSLABTEST OFCONCRETE

A sample used for the experiment and formonitoring of foundation – subsoil interaction was asteel-fibre reinforced concrete slab. The size of thefibre concrete slab was 2000 x 2000 x 170 mm. The
C25/30 concrete was cast there - it was reinforced
with scattered reinforcement. The reinforcement
consisted of steel fibres, 3D DRAMIX 65/60B6–25
kg.m-3. The slab casting process is shown in Fig. 1.

Fig. 1 Casting of the steel-fibre reinforced concrete
slab

473

GEOMATE- Bangkok, Nov. 14-16, 2016is different from that in the homogeneous half-space.
The modulus of deformability changes continuously,
depending on the depth. This means that it is themodel of the inhomogeneous half-space which
describes the deformation behaviour of the soil
better that the homogeneous model. Using of
inhomogeneous half-space is also described in
papers [12, 13, 14]. Homogeneous half-space andalso inhomogeneous half-space were used innumerical analyses.

The upper layer of subsoil consists of loess loamwith F4 consistency. Thickness of that layer is about
5 meters. Poisson coefficient of the subsoil isµ = 0.35, and modulus of deformability is E DEF =23.7 MPa. From the geologic point of view,foundation soil is not complex.
During the test, the fibre concrete slab wasloaded in the middle by the pressure applied by thehydraulic press. Dimensions of the area under load
were 200 x 200 mm. The loading was carried out insteps: 50 kN / 30 minutes. The slab failed during the
6th cycle when the loading force was 250 kN. The
model of steel-fibre reinforced concrete foundationslab was violated by punching shear. Punching shearof slabs is discussed in detail [1].

Fig. 2 Load test of the steel-fibre concrete slab
3D COMPUTATIONAL MODEL IN ANSYSSYSTEM
A 2D element, SHELL 181, was chosen for thefoundation slab which was modelled as a surface
with the specified slab thickness. The subsoil wasmodelled using a 3D element, SOLID 45. The
material properties were modelled using themodulus of elasticity E (or the modulus ofdeformability E DEF in case of soil) and Poissoncoefficient µ.
In connection with the creation of a spatial modelusing 3D elements is particularly problematic
correctly to determine the size of the modeled area
representing the subsoil, to choose boundary
conditions and finite size mesh.
Soil is inhomogeneous and its properties aredifferent from idealised properties of an elastic
isotropic and homogeneous substance. In thehomogeneous half-space, the modulus ofdeformability is constant and does not depend on
depth (Fig. 3). In the inhomogeneous half-space, theconcentration of vertical stress in the foundation axis

Fig. 3 3D numerical model in ANSYS
Shell 181 is a four-node element with six degrees
of freedom in a node. The degrees of freedom
represent three dislocations in x, y and z axes andthree torsional displacements around x, y and z axes.
SOLID 45 is defined by eight nodes where eachnode is characterised by three degrees of freedom(dislocations of the node in x, y and z axes). Deadweight of soil and dead weight of concrete slab were
not considered for the calculation. The slab andsubsoil were modelled using the regular finiteelement mesh. The sizes of mesh elements were474

GEOMATE- Bangkok, Nov. 14-16, 2016

different for the subsoil and for the slab surface
where the mesh was denser. The force was specified
for each node in the finite-element mesh of the slab.
The location and value corresponded to the load
applied onto the slab during the experiment. In order
to transfer effects of the load, which was applied on
the foundation slab, into the subsoil it was essential
to create a mutual contact and define a contact
surface. The FEM model was solved using thecontact elements.
Contact elements mediate the kinematic processof deformation. The contact was created using thecontact pair: TARGE170 – CONTA173. Friction
between the slab and subsoil was neglected there.

Fig. 5 Variant of boundary condition C
Deformation versus variable depth of the subsoil
modelFig. 6 - Fig. 8 compares the deformations of theslab for the variable depth of the homogeneous halfspace (light curve) and inhomogeneous half-space(dark curve) for all variants of the boundary
conditions.
The ground plan is 6.0 x 6.0 m and boundary
conditions are A:

PARAMETRIC STUDY
In connection with the creation of a spatial modelusing 3D elements is particularly problematic
correctly to determine the size of the modeled area
representing the subsoil, to choose boundary
conditions and finite size mesh. A dependence
deformation on these parameters was proved byparametric study. Two aspects were consideredwhen comparing different models. One aspect isdependence of deformation on variable depth of the
subsoil, while keeping the same ground plan of the
subsoil. The second aspect is dependence of
deformation on variable size of ground plan of the
subsoil, while keeping the same depth. The
comparison was made for three different boundary
conditions - see Fig. 4, Fig. 5.

Fig. 6 Deformation versus variable depth of the
subsoil model, boundary conditions A
The ground plan is 6.0 x 6.0 m and boundary
conditions are B:

Fig. 7 Deformation versus variable depth of the
subsoil model, boundary conditions B
Fig. 4 Variant of boundary condition A, B

475

GEOMATE- Bangkok, Nov. 14-16, 2016

The ground plan is 6.0 x 6.0 m and boundary
conditions are C:

The depth is 6.0 m and boundary conditions C:

Fig. 11 Deformation versus variable ground plan of
the subsoil, boundary conditions C

Fig. 8 Deformation versus variable depth of the
subsoil model, boundary conditions C

The deformation in the middle of the steel-fibre
reinforced concrete slab measured during theexperiment was ca. 2.83 mm.

Deformation versus variable ground plan of the
subsoilFig. 9 - Fig. 11 compares the deformations of theslab for the variable ground plan of thehomogeneous half-space (light curve) andinhomogeneous half-space (dark curve) for all
variants of the boundary conditions.
The depth is 6.0 m and boundary conditions A:

CONCLUSIONS
It follows from the characteristics deformation vs.
variable depth of the subsoil model that theincreasing depth results in increases in the verticaldeformation. Considering different variants of theboundary conditions, one can compare the growth ofvertical deformation once the depth becomes higher(Fig. 12).

Fig. 9 Deformation versus variable ground plan of
the subsoil, boundary conditions A
The depth is 6.0 m and boundary conditions B:

Fig. 12 Deformation versus variable depth of the
subsoil model, all boundary conditions
In case of the variant C when all walls except for
the upper surface are fixed, the boundary conditions
play such an important role that even the increasing
depth does not increase the vertical deformation somuch as for other variants of boundary conditions.
On the other hand, the least influence of boundary
condition was for the variant A, when deformation isFig. 10 Deformation versus variable ground plan of
the subsoil, boundary conditions B

476

GEOMATE- Bangkok, Nov. 14-16, 2016homogeneous half-space. This is really great
variance in the resulting deformation depending on
the size of the subsoil model and its boundary
conditions. Fig. 12 and Fig 13 the verticaldeformation is ranges from 2.207 to 2.759 mm inanalyses of inhomogeneous half-space. This is not
so really great variance in the resulting deformation
depending on the size of the subsoil model and its
boundary conditions as in analyses of homogeneous
half-space. This means that the inhomogeneouscontinuum provides more stable results which areless affected by the choice of the geometry anddimensions of the area representing the subsoil.

possible for each wall of the subsoil model in eachdirection except for the lower base which is fixed.
Fig. 12 shows that the higher the depth of the subsoil
model is, the bigger the difference is between
deformations calculated for the variants of boundary
conditions. With the increasing depth of the subsoil
model, the selection of boundary conditions is
becoming a more important criterion which
influences the final vertical deformation.
An important lesson learnt from characteristics
describing the deformation versus variable size of
subsoil ground plan is that the influence of any
boundary condition is becoming weaker with theincreasing ground plan of the subsoil. Using thechart in Fig. 13 it can be concluded that theboundary conditions play no role at all, if the groundplan of the subsoil model is big enough.

ACKNOWLEDGEMENTS
This outcome has been achieved with thefinancial support of the project GACR No. 1608937S „State of stress and strain of fiber reinforced
composites in interaction with the soil environment“.
In this undertaking, theoretical results gained byconceptual development of research, development
and innovations for 2016 at the VŠB-Technical
University of Ostrava (granted by the Ministry of
Education, Youths and Sports of the Czech
Republic) were partially exploited.
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ISSN 1877 - 7058,[2] M. Aboutalebi, A. Alani, J. Rizzuto, D. Beckett,“Structural behaviour and deformation patterns
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[6] A. Alani, M. Aboutalebi, “Analysis of thesubgrade stiffness effect on the behaviour ofground-supported concrete slabs” StructuralFig. 13 Deformation versus variable ground plan of
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has been proved by the chart in Fig. 13.
The charts prove the major role played by thesize of the model area and by the boundary condition
itself. Results were quite scattered for the variants.
Charts in Fig. 12 and 13 indicate how the parametersof the area influence considerably the resulting
vertical deformation.
From deformations obtained by numericalsubsoil model of homogeneous half-space, theconclusion can be drawn that the values ofdeformation resulting from the 3D model are too
scattered. Final deformation measured during theexperiment was ca. 2.83 mm. In this set of results in
Fig. 12 and Fig 13 the vertical deformation is ranges
from 2.926 to 4.708 mm in analyses of477

GEOMATE- Bangkok, Nov. 14-16, 2016

Concrete, Issue 13, Volume 2, pages 102-108,
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[10] R. Cajka, P. Mateckova, ”Comparison of
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Slabs on the Subsoil”, Transactions of the VŠB– Technical University of Ostrava, Civil
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[16] R. Cajka, “Analytical derivation of friction
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[18] J. Labudkova, R. Cajka, “Comparsion of
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[19] Cajka, K. Burkovic, V. Buchta, R. Fojtik,“Experimental soil - Concrete plate interaction
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[20] Bin Tian, Rui Li, Yang Zhong, “Integral
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478

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INVESTIGATION OF POTENTIAL ALKALI-SILICA REACTIVITY OF
AGGREGATE SOURCES IN THAILAND

Suvimol Sujjavanich1, Krit Won-In2, Thanawat Meesak1, Watcharagon Wongkamjan3, Viggo Jensen4
1
Faculty of Engineering, Kasetsart University, Thailand
2
Faculty of Science, Kasetsart University, Thailand
3
Faculty of Liberal Arts and Sciences, Roi Et Rajabhat University, Thailand
4
Norwegian Concrete and Aggregate Laboratory LTD, Norway

ABSTRACT
Since the first local evidence of alkali-silica reaction (ASR) was reported in Thailand in 2011, awareness of
ASR was raised and the importance of aggregate performance database was recognized, thus an attempt to set up
local aggregate database was initiated. As part of the effort to establish the database, a study of potential ASR
reactivities was carried out involving aggregates from several current industrial sources in eastern and central
Thailand and the results were presented here. Aggregate samples were randomly collected from various sources
and tested by two methods; accelerated mortar-bar test (AMBT) and petrographic examination. The types ofaggregate sampled included limestone, greywacke, and rhyolitic tuff. The AMBT expansion results indicatedthat several aggregate types had larger expansion than the threshold of potentially deleterious ASR behavior assuggested by ASTM C 1260 standard. Furthermore, it was found that samples of same aggregate type, althoughapparently had mineralogically similar compositions, had different reactivities, particularly when sampled fromaround geological fault zone compared with that taken from the surrounding area and portions of parent rockwere observed to be weathered in this area. Thin section analysis revealed evidence of ASR gel at the aggregates’
rim, inside the aggregates and in the matrix. These findings suggest possible future ASR problems in Thailand as
well as in neighboring countries where the continuity of geology pattern is observed.

Keywords: Aggregate database, Alkali silica reaction (ASR), Expansion, Petrographic Analysis, Thin Section

Thailand is situated on two adjacent terranes, theShan-Thai and the Indochina terranes. The ShanThai terrane covers the east of part of Myanmar, thewest, north, central and southern parts of Thailand
with rocks dating back to the Precambrian eon. The
Indochina terrane covers the northeastern andeastern parts of Thailand as well as Cambodia andthe majority of Laos and Vietnam landmasses withrocks dating back to Paleozoic era. The suturejoining the two run down north – south direction
giving rise to the Loei-Petchabun mountain ranges
that stretches from Thailand’s northern border with
Laos in Loei province down along the Petchabun
fold belt bordering the rim of the Khorat plateau to
the east and the central plain to the west. Evidenceof ancient volcanic activities can be found along thePetchabun fold belt and igneous rocks such as
granite, granodiorite, diorite, monzonite, rhyolitic
tuff, andesite, and tuff. The suture runs further down
through the eastern region of the country andextends to the upper part of the Gulf of Thailand.

INTRODUCTION
The first evidence of alkali-silica reaction (ASR)
in Thailand was found in one large infrastructure in2009 and reported by Sajjavanich in 2011 [1]. The
confirmed problematic rocks were black quartzite
with traces of pyrite, mica and sericite. Some ofthese rocks were confirmed as slow late reactive
types [2]. Therefore, the awareness of ASR was
raised and the importance of aggregate’s
performance database was recognized as a basis forachieving a better understanding of the reactivity of
local materials and the consequent activities to cope
the future problems including mitigation, both forresearch and construction practice. Thus an attempt
to set up local aggregate database was initiated. As
one part of the ongoing effort to establish the
database, a study of potential ASR reactivities was
carried out involving aggregates from several
current industrial sources in eastern and central
Thailand.

479

GEOMATE- Bangkok, Nov. 14-16, 2016

This study investigated aggregates from working
industrial mining operation from the central andeastern region of Thailand. These two regions ofinterest in Thailand have significant differences in
landforms. The eastern region is bordered by the
Gulf of Thailand to the south and southwest, thesouthern part of Loei-Petchabun mountain ranges tothe northwest and the rim of the Khorat plateau to
the north. The area includes undulating coastal
plains between the sea and the mountains covering
the provinces of Chonburi, Rayong, Chantaburi andTrat. The landform in this region is highly
discontinued with evidence of some geologic
activities thus rocks with high degree of alteration
were not uncommon in these areas. Granite can befound in three separate areas, in Chonburi, Rayong,and Chantaburi, and rhyolite can be found in theextreme east of the region close to the border with
Cambodia. High-grade metamorphic rocks with
important rock-forming minerals such as quartz,
biotite and feldspar can also be found in Chantaburi
and Trat provinces. Various sedimentary rocks arefound in this region such as quartzite, shale, chert,
and limestone found along Chonburi’s coastline.
Limestone, shale and greywacke are also found in
Chantaburi and Trat provinces. Moreover, becauseof the suture running through this area and activities
of various local faults, the characteristics of rocks
found in this area are quite complicated. Some rocks
have undergone metamorphism, others haveundergone alteration, and the effect can be very
localized. History of geological landform and timestrongly influence the changes of some of these to
low-grade metamorphic rocks [3]. The dominantcharacteristics of discontinued landform explain thevariations of type of rock, which are the sources ofindustrial aggregates in these areas.
The central region of Thailand consists of fluvial
flood plain bordering highlands and mountainous
regions to the north and west, flanked by the LoeiPetchabun mountain ranges to the east and extending
south to the Gulf of Thailand. The country’s majorrivers flow through this region depositing sedimentfrom the highlands. Coastal deposit is found in thelower third of the region. The upper central plain
region consists of Uttradit, Sukhothai, Pitsanulok,
Pichit, and Kampangpet provinces and northern partof Nakhonsawan province. These provinces lie inthe confluence of the major rivers that join together
further south in Nakhonsawan province to form
Chao Phraya river which forms the border of upperand lower central plains. The lower central plain
begins in the southern part of Nakhonsawanprovince and continues south until the mouth of the
Chao Phraya river on the gulf of Thailand. There are
isolated rock formations scattered around the region
with formations of plutonic rocks such as granite,
granodiorite and diorite in the south and southwestof Nakhonsawan and formations of volcanic rocks

such as rhyolite, andesite, and dacite in the east of
the region as forming part of the Saraburi group ofPermian rock formations [4]. In Saraburi area, there
are six known formations with mixed carbonate andclastic sequence of outcrops, especially on the mosteastern area bordering the western rim of the Khorat
plateau [5]. This area is also the source of goodquality limestone used in construction industry andcement industry. Gray shale and siltstone with minor
or major of sandstone and limestone are found insome formations while others may be characterizedby limestone with interbedding with dolomitic
limestone, shale, sandstone and tuffaceous sandstone
including rhyolite dikes and chert nodules. In
addition to the good quality limestone, which
typically has been used for construction in the past,
it is of interest that the diversity of landform andgeological nature indicates various natural rocktypes in these two areas and these have been utilized
as natural resources for construction and otherindustries. Some of them have been widely reported
as potentially reactive aggregates [6]. The variationsof these geological parent rocks have strong impact
on the local industrial aggregates, which is the main
component of concrete, and the chemical
deterioration of concrete structure may be aconsequence if the potential risk of reactivity of therock is not properly assessed.
All of the aggregate mines are either privately
owned or have been granted long-term government
mining concessions and most of them are used for
concrete production. Since no concrete deterioration
problems as a result of reactive aggregates had
previously been reported, it was not uncommon that
there was a lack of local information concerning theaggregates except the conventional geology
classification. Therefore this paper reports the resultsof an investigation of the industrial aggregates interm of both geological and chemical durability inengineering aspects. This is expected to provide
information and insight into potential risk ofchemical deterioration in future concrete structuresin Thailand and neighboring countries.
MATERIALS AND METHODS
Materials
Three common rock types used as aggregates inThailand were selected for this study, namely,
rhyolitic tuff (RYSR) from the central region,
limestone (LSCB) and greywacke from the easternregion. The greywacke aggregates were from two
different sources approximately 100 km apart, one in
Chantaburi province (GWCT), and the other in Trat
province with two samples taken from this quarry(GRA and REA)
The three rock types were picked because of480

GEOMATE- Bangkok, Nov. 14-16, 2016

RESULTS AND DISCUSSION

their differences. While the major mineralcompositionsoflimestone,anon-clastic
sedimentary rock, are calcite and aragonite which
are only different crystallographic forms of calcium
carbonate (CaCO 3 ), on the other hand, greywacke, aclastic sedimentary rock, is a result of sedimentationand lithification of mainly sandstone and other small
rock fragments, angular quartz grains, feldspar
crystals in a matrix of fine clay, whereas, rhyolite, afelsic volcanic rock is an igneous rock rich in silica.
These three different rock types were
investigated in details for microstructure, mineraland chemical compositions as well as expansion in
the accelerated mortar bars.
Type I Portland cement, according to ASTM
C150-07 [7], with Na 2 O equiv. of 0.409 % was used
for this study.

Microstructure Analysis
The XRF results of all the aggregates are shownin Table 1. The investigation showed the major
compositions of all studied aggregates. All thegreywacke samples even though came from two
different sources (GRA and REA from one, GWCTfrom the other) had very similar composition withamounts of major oxides falling in the same ranges(such as SiO 2 56-58%, Al 2 O 3 13-18% and Fe 2 O 3 57%, CaO 5-7%, MgO 1.9-3.2%). However, thepetrographic analysis and XRD analysis were
performed on the same specimens and the results
showed significant difference, especially the
microstructure of GRA and REA observed in thinsection petrographic analysis. GRA aggregates,
which were collected from the same mine as REA
but from the general areas of mine was coarse-grain
greywacke. The microstructure of GRA is shown in
Fig. 1 (a) composing of quartz, feldspar, and rockfragments. This was similar to the microstructure of
GWCT aggregates that were collected from the otherquarry. However, the REA samples, Fig. 1 (b), werecollected from the same quarry as GRA but from theshear zone showed fine-grained matrix ofsubgrained quartz and rock fragments ofclay/graphiteandhigh-weathering
chlorite,
deformed by tectonic forces. The mineralogicalcompositions were confirmed by XRD analysis.
Quantitative XRD analysis showed high content of
muscovite and low content of albite found in REA,
31.9% and 7.1% respectively, which was opposite ofwhat was found in GRA where low amount of
muscovite and high amount of albite was found,
13.1% and 32.1% respectively. Another striking
difference was the difference in granularity of
crystals in REA and GRA. REA consisted of small
grains of quartz and rock fragments instead of big
grains such as those found in GRA. The
microstructure of REA consisted of deformationinduced subgrained quartz particles to become

Methods
To investigate the potential alkali-silica
reactivity of the various aggregate types, samples ofaggregates were taken from different sources inmany parts of the country and were tested bylaboratory
methods,
namely,petrographic
examination and accelerated mortar-bar test
(AMBT).
Thin section petrography was carried out using
Olympus BX-41 imaging petrographic microscope
equipped with plane and cross polarization
illumination mode. The petrographic analysis wasused to study the microstructure of the aggregates, toidentify the constituent minerals and to classify theaggregates. The chemical compositions wereanalyzed with X-ray Fluorescence spectroscopy(XRF) and X-ray diffraction spectroscopy (XRD).
Accelerated mortar-bar test (AMBT) was used tomeasure expansion due to alkali silica reactivity of
the aggregates and was carried out according to
ASTM C 1260 [8] standard testing procedure.

Table 1 Chemical composition of materials used in this study
SiO 2

Al 2 O 3

Fe 2 O 3

CaO

MgO

SO 3

Na 2 O

K2O

LOI

GRA

58.22

13.93

6.11

6.96

2.53

0.24

3.08

1.32

6.77

REA

56.48

18.64

5.29

3.21

1.9

0.38

1.15

3.39

8.36

GWCT

57.63

15.33

7.37

5.52

3.19

0.24

2.97

1.41

5.31

RYSR

56.49

17.39

6.51

8.55

3.02

0.04

2.69

4.03

na

LSCB

34.24

7.99

4.53

25.94

2.25

1.92

0.42

2.47

19.34

Cement

18.74

5.22

3.2

65.3

0.82

2.8

0.08

0.5

2.75

481

GEOMATE- Bangkok, Nov. 14-16, 2016

a

b

2 mm

2 mm
c

d

2 mm

2 mm

Fig. 1 Thin section microscopy of different aggregates studied (a) greywacke (GRA), (b) greywacke(REA), (c) limestone (LSCB), and (d) rhyolitic tuff (RYSR)
microcrystalline/cryptocrystalline matrix mixed with30% clay/graphite matrix as shown in Fig. 1 (b).
This may explain the highest ASR reactivity in thisaggregate group and the proposed classification as“very likely to be alkali-reactive.
Compared to those of greywacke, the chemicalcompositions of rhyolitic tuff appeared to fall in thesame ranges, only a slightly high oxide content ofcalcium and magnesium and low SO 3 were observed.
Slight weathering was observed in hand specimen
and from the microstructure analysis. The mineralcompositions of the very fine-grained, massive andslightly weathered pale green-gray colored sample
was tested using XRD analysis and was found to
consist of albite, quartz, calcite, chlorite, K-feldspar
and illite. The mineral compositions content frommodal analysis indicated subhedral crystals ofplagioclase and pyroxene of about 70% and 10%
respectively. The anhedral crystals of opaqueminerals and calcite were 10% and 5% as well as
small amount of olivine, chlorite and epidote ofabout 1-2% as minor minerals. The average
phenocrysts consisted of 0.05-3mm in size and mostplagioclase showed zonal texture and lath shape.
Some glassy phases were also observed, which,
together with microcrystalline and cryptocrystalline
quartz present, may account for its reactivity. Other
authors also found this rock type to be a reactive
aggregate in many countries [2,6].
The non-clastic and stylolitic sedimentary rock,limestone, from the eastern province of Chonburi
was slightly high in silica (34.24%) but moderate in

CaO (25.94%) with relatively high LOI (19.24%).
Because of the high silica content, this aggregate canbe classified as siliceous limestone and other authors
found this rock type to be reactive [9]. The mineral
grain size of the light gray colored sample ranges of
0.1-2 mm, while modal analysis showed major
minerals to be calcite, clay, and microcrystalline
quartz with some calcite veinlet and quartz veinlet.
Opaque minerals, possibly pyrite, were also found in
the thin section, which would account for its high
LOI. No weathering but fracturing was morecommonly observed.
AMBT Results
The results of the AMBT expansion tests of all
the studied aggregates, limestone, greywacke andrhyolitic tuff, are shown in Fig. 2. The average
expansion of each aggregates were higher than thesuggested limit of 0.2% at 14 days exposure to
sodium hydroxide solution (except rhyolitic tuff
which was slightly less), which indicated that all
three rock types were potentially reactive aggregates.
The deformation induced subgrained greywacke(REA), showed the highest expansion and rhyolitic
tuff the lowest. The comparison between the average
expansion of greywacke with different granularity
from different location of the same mine (REA andGRA) and the average expansion of similar
aggregate type from different mine (GWCT) werealso demonstrated in Fig. 2. These agreed well withthe petrographic analysis. The effect of deformation482

GEOMATE- Bangkok, Nov. 14-16, 2016

0.350
0.300
0.250

% Expansion

0.200
0.150

RYSR

0.100

LSCB
GWCT

0.050

GRA

0.000
0
-0.050

2

4

6

8

10

12

14

16

REA

Days after castingFig. 2 AMBT results of aggregates. Different granularity of greywacke aggregates affects their average
expansions under accelerated conditions
induced subgraining resulted in REA yielded thehighest expansion than those of aggregates fromtypical area (GRA) and from other mine (GWCT). It
was clear that mineralogical and the past history of
the studied three types of aggregates were strongly
influenced the reactivities of these materials, not thechemical compositions.
The expansion results indicated that all studied
aggregates, whatever types, had larger expansion
than the threshold for potentially deleterious ASR
behavior as suggested by ASTM C1260 standard.
Furthermore, samples of the same aggregate type,although with mineralogically similar compositions,
had different reactivities, particularly when sampledfrom around geological fault zone compared withthat taken from the surrounding area. Portions ofparent rocks in this area were observed to be
weathered. In addition, thin section analysis ofconcrete prisms made from REA and GRA fromprevious study [10] revealed evidence of ASR gel at
the aggregates’ rim, inside the aggregates and in thematrix. In light of these findings, it suggests possible
future ASR problems in Thailand as well as inneighboring countries where the continuity ofgeology pattern from Thailand or similar are
observed.

some sources are used, in combination with somespecific conditions such as high humidity
environment. This is also important for theneighboring countries such as Laos, Myanmar andCambodia, in which there has been no AAR
deterioration reported yet. The similar condition
both for geology, materials and climatic condition isof concern for the potential AAR in this region.
CONCLUSIONS
From the study, conclusions as follows are−Two types of sedimentary rocks, greywacke
and limestone from some sources in eastern regionof Thailand, as well as an igneous rock, rhyolitic tuff,
from the central region, showed the potential
reactivity as the root of ASR problem in the future.
−Greywacke from the different sources were
almost similar in compositions, but significantly
different features have been observed in thin-section
analysis. GRA and CTGW were coarse-grained
greywacke composed of quartz, feldspar, and rockfragments but REA had different granularity than theother two, with high-weathering chlorite with amatrix of subgrained quartz, rock fragments andclay/graphite, deformed by tectonic forces. The ASR
risk is increased significantly from subgraining of
the microstructure. The highest expansion of REA
agreed well with the microstructure analysis.
−The microstructure investigation of volcanic
rock aggregate, rhyolitic tuff showed slightweathering and the expansion result indicated

The ASR Risk In Thailand And South East Asia
The hot and humid weather in Thailand is typicalfor this region. This may create high risk of ASR in
the future if the potentially reactive aggregates from483

GEOMATE- Bangkok, Nov. 14-16, 2016

potential reactivity.
−The high silica content found in the limestone
sample studied and the expansion higher than thelimit suggested potential alkali reactivity.
−All studied samples in this paper revealed thepotential alkali reactivity although aggregates types
are different.
At present, ASR is not a serious problem inThailand and only one case is reported. Howeverawareness that ASR might be a possible problem in
the near future is important based on the varieties ofaggregates from local sources as well as thechemical variation particularly the alkali content oflocal cement. Therefore, continuing work on
database development may be an appropriate use ofavailable resources.

Aggregate Reaction in Concrete London, 1992.
Vol. 1, pp. 477-484.
[3] DepartmentofMineralResources(2001):Geology of Thailand, 5th December 1999
HM the King Golden Jubilee Commemoration
Edition, Department of Mineral Resources in
2001[4] Bunopas, S (1981): Paleogeographic history ofwestern Thailand and adjacent parts ofSoutheast Asia a plate tectonics interpretation,
Ph.D. Thesis, Victoria University of Wellington,
New Zealand, Reprinted 1982 as Geological
Survey Paper No. 5, Department of Mineral
Resources, Ministry of Industry, Thailand, 810[5] Hinthong, C (1985): Upper Paleozoic system:
the central plain and lower Phetchabun ranges.
Geological Survey Division, Bangkok, 42 pp.
[6] R.N.Swamy, ed. (2003): The Alkali-Silica
Reaction in Concrete: Taylor & Francis eLibrary, 2003.
[7] ASTM C1260-01 (2001): Standard Test Method
for Potential Alkali Reactivity of Aggregates(Mortar-Bar
Method),
ASTM
International, West Conshohocken, PA, 2001,
www.astm.org[8] ASTM C150-07 (2007); Standard Specification

ACKNOWLEDGEMENT
The authors would like to thank Kasetsart
University Research and Development Institute(KURDI) and National Research Council ofThailand (NRCT) for their financial support for thisstudy.
REFERENCES[1] Sujjavanich,S,

Suwanvitaya,

S,

for Portland Cement, ASTM International, WestConshohocken, PA, 2007, www.astm.org.
[9] Y. Monnin, P. Dégrugilliers, D. Bulteel, and E.
Garcia-Diaz (2006): Petrography study of two
siliceous limestones submitted to alkali-silica
reaction. Cement and Concrete Research, vol.
36, no. 8, pp. 1460–1466, Aug. 2006.
[10] Wongkamjan, W, Sujjavanich, S, and Won-In,andRothstein, D (2011): Investigation of ASR in
Mass Concrete Structures: A case study.
Modernmethods
andAdvances
in
Structural Engineering and Construction(ISEC-6), Zürich Swizzerland 2011, pp.
1297-1302. Research Publishing Services.
[2] Jensen V, and Danielsen S. W (1992): AlkaliAggregate Reaction in Southern Norway. The
Ninth International Conference on Alkali-

K (2014): Influence of Recrystallized Silica
Aggregateson Alkali-Silica Reactivity.
Kasetsart Journal: Natural Science 48(3).

484SixthInternational Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand,Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

CBR BEHAVIOR OF SOFT MARINE CLAY TREATED WITH CLASS‘F’ FLY ASH AND NANO MATERIAL
Prasanna P.Kulkarni 1 and J. N. Mandal2
1

2

Department of Civil Engineering, Indian Institute of Technology Bombay, Mumbai, India
Faculty, Department of Civil Engineering, Indian Institute of Technology Bombay, Mumbai, India.

ABSTRACT
Soil improvement in expansive soils is an exigent task and it can pose a lot of constructional problems if not
tackled in a proper way. Also, the disposal of fly ash, available in abundance in the vicinity of thermal power
stations is the biggest global concern today from the environmental perspective. An attempt to improve the soft
marine clay soil CBR properties by adopting class ‘F’ fly ash and organosilane nanomaterial was made in thisresearch by conducting modified CBR tests. Marine clay was treated with Class ‘F’ fly ash with 0, 5%, 10%,15% and 20% by dry weight of soil. Each proportion was further treated with nano material diluted in water withfour different dilution ratios of (1:150), (1:300), (1:600) and (1:800) by volume. An improvement of 188.88% in
the soaked CBR was observed on account of addition of 10% of fly ash to the soil. Further, it was observed thatsoaked CBR values were drastically improved on treating the soil with nano material. Maximum improvements
were observed in soil treated with 10% fly ash and nano material with dilution ratios of (1:150) and (1:300).
With 10% flyash and dilution ratio of (1:150) the soaked CBR value was observed to be enhanced by 381.66%
and with (1:300) dilution ratio the soaked CBR was enhanced by 478.33%, compared to the neat soil. Also thevalues of Atterberg’s limits were observed to be considerably decreased on account of treating the soil with 10%
Fly ash and nano material of dilution ratios (1:150) and (1:300). This research is thus an attempt to stabilize thesoil by utilizing the fly ash along with innovative nano-technology which can provide an eco-friendly solution
and engender a revolutionary change in pavement infrastructural applications.
Keywords: Marine Clay, Class ‘F’ Fly Ash, Nano Material, California Bearing Ratio, Atterberg’s Limits
carried out to make the best use of this rich siliceous
material to cater various needs in engineering andtechnology.
Also, there is one such upcoming field of nanosciences which has brought revolutionary changes in
the field of science, engineering and technology
since last decade. Organo silane compound is one
such promising material which works on the conceptof nano technology and promises to be very effectivein stabilization of soft soils particularly for the roadapplications.
The study aims at optimizing the proportions ofClass ‘F’ fly ash in suitable combinations with nanomaterial and soft soil to unveil the improvements inthe engineering parameters which will cater theneeds of good subgrade material for pavements.

INTRODUCTION
The overall socio-economic growth of a developing
nation highly depends upon the pavement
infrastructure of the country. The soil improvement
in expansive soils is an exigent task and it can pose a
lot of constructional problems like swelling,
shrinkage etc. if not treated properly. Therefore
stabilization of expansive subgrades becomesmandatory to enhance the life of pavement
infrastructures. Many researchers earlier have tried
to stabilize the expansive soils by fly ash.
In developing countries like India, theexpeditious industrialization has led to more power
generation demands. More and more power
generation initiates greater production of fly ash.
The fly ash production in India is expected to reachby 225 million tons by 2017 [1]. Huge quantity of
fly ash lying in the vicinity of thermal power stations
needs quicker disposal and utility. If not properly
disposed it may cause severe health hazards to the
mankind. Therefore the problem of disposal of fly
ash needs to be very skillfully dealt with from theenvironmental perspective. In the recent years
utilizing fly ash as an alternative construction
material for many important projects has become atrend of the industry and lot of research is being

PRIOR ART
Many earlier researchers have tried to assess thesuitability of fly ash in various applications since
past two decades. Studies on soil stabilization withfly ash and other cementitious products have been
carried out to focus on the effective utilization of fly
ash as a suitable material for stabilization of soft
soils and subgrades. Pulp mill fly ash and lime has
been utilised for improvement in bearing capacity
485

GEOMATE-Bangkok, Nov. 14-16, 2016[2]-[3]. Studies on use of class ‘F’ fly ash with soilcement and soil-lime for base layers in highways
have been presented as in [4]. The improvements ingeotechnical properties of expansive soils byadopting varying percentages of fly ash have beenpresented as in [5]. Saturated drained triaxial tests
were conducted to assess the behaviour of soils withaddition of fly ash and calcium chloride considering
the curing conditions as in [6]. Randomly oriented
polyester fiber inclusions with fly ash and cement
additive combinations were studied to bring moreductility to the soil mixes [7]-[8].
In-situ case studies on fly ash stabilized soils also
have been undertaken to measure the increase instrength and stiffness of fine subgrade by measuring
the CBR and resilient modulus properties [9]. Waste
plastic strips have been used in combination withstone dust and fly ash to observe the effect on CBRand secant modulus values as in [10]. Sewage sludge
ash has been effectively established as a stabilizing
material with lime in improving strength, CBR andplasticity characteristics of soft soil [11]. The
influence of fly ash on UCS of soil by placing soil
and fly ash layers successively has been presented asin [12].Studies on basic characterization of class ‘F’
fly ash to be utilized as backfill material inreinforced earth walls have been presented as in [13].
Studies on behavior of cellular reinforced earth wallswith fly ash as backfill under strip loading
conditions have been discussed as in [14].
Improvement in CBR behavior in soil adopting class‘F’ fly ash in varying percentages has been presented
as in [15]. Studies on strength and compressibility
improvement in marine clay by treating with lime
for strong reliable roads and buildings have beenpresented as in [16]. Stabilization of expansive black
cotton soil by dolime fines for utilization in subbasesof pavements have been discussed as in [17].
The importance of nano materials grew over the
past decade and thereafter few researchers have tried
to ascertain the versatility of these very novel
materials in various streams of science andtechnology. Few researchers have presented theimportance of nano technology in constructions.
Nano-silica was found to be effective in increasing
the pozzolanic activities of the high volume fly ash
high strength concrete [18]. Strength properties ofsewage sludge ash were improved by adding nanosilica and nano alumina additives to soft soils insmall varying proportions [19]-[21]. Use of nanoaluminium has been reported to study the transport
through the porous media with the introduction incolumn studies to measure the leachate
concentrations [22]. The physical, mechanical andthermal properties of concrete were improved byaddition of nano zinc oxide as in [23]-[24]. Studies
on improvement in CBR values on Nigerian soils bytreating them with nano material have beenpresented as in [25]. Studies on improvement instrength characteristics of silty sand samples byadding colloidal nano silica have been discussed asin [26].
Earlier researches report the several applicationsof nano materials in almost all fields of science andengineering. However it is seen that major work is
been done related to cement and concrete industry.
However there is very less evidence of work related
to the field of geotechnical engineering andpavements and yet there seems to be ample space forresearch to showcase the advantages of nanomaterials in soil sciences. This research is thus anattempt to glorify the prospective future ofnanotechnology in construction by utilizing fly ash,which will enable in solving the disposal problem toa certain extent.
CHARACTERIZATION OF MATERIALS
Fly ashFly ash was procured from a thermal power stationnear Khopoli, in the Raigad district of Maharashtra.
The fly ash is being denoted here as ‘FA’. Table 1
represents the physical properties of fly ash.
Table 1 Physical properties of Fly ash and soil
Properties
Specific gravity
Maximum dry density (gm/cm3)
Optimum moisture content (OMC) %Coefficient of permeability, k(cm/sec)
Gravel
Sand
Silt
Clay
D 10 mm
D 30 mm
D 60 mm
Uniformity coefficient (Cu)
Coefficient of curvature (Cc)Group symbol
Liquid Limit (%)
Plastic Limit (%)
Soaked CBR (%)

Values
2.093
1.07
28.5
1.061 x 10-6
-26.5%
70%
3.5%
0.007
0.02
0.046
6.57
1.24
SM
Non Plastic
Non Plastic
5.6

Chemical analysis of Fly ash was carried out by XRay Fluorescence test (XRF). Table 2 depicts thechemical composition of Fly ash. The Fly ash is seen
to be very rich in Silica (SiO 2 ) of about 62.7% and a
low Calcium oxide content (CaO) of 0.905%. The
Fly ash was classified as Class ‘F’ Fly ash as per[27].
Table 2 represents the chemical composition of fly
ash.

486

GEOMATE-Bangkok, Nov. 14-16, 2016

Table 2 Chemical composition of Fly ash

Nano material

Constituent
(%)
CaO
0.905
Fe 2 O 3
5.152
K2O
0.927
MnO
0.029
P2O5
0.188
SO 3
0.015
SrO
0.056
TiO 2
1.642
Al 2 O 3
26.986
BaO
0.068
MgO
0.545
SiO 2
62.799Na 2 O
0.110
Fig. 1 represents the electron microscopic image of
Fly ash obtained by Field emission Gun Scanningelectron Microscope (FEG-SEM).

Nano material is basically a organo silane compound
and was obtained from Zydex Industries, Vadodara,
Gujarat, India. The use of commercial name of thematerial has been avoided to make it more palatable.
The organo silane compound used in the research
is a pale yellowish liquid which could be dissolvedin water. The compound on reacting with water i.e.
hydrolysis forms strong bonds when the nanosolution with water is mixed with the soil. The
solution coats the soil surfaces by converting water
absorbing groups to water repellent surfaces [25].
Since silica has a great affinity towards oxygen,string bonds are formed which create a thinhydrophobic film at the molecular level treatedsurfaces which makes the soil impermeable thus
causing a significant improvement in its
geotechnical properties.
Table 4 represents the CHNS analysis of nanomaterial.
Table 4 CHNS Analysis of Nano material
Constituent
Percentage (%)
Carbon64.173
Hydrogen9.669
Nitrogen
3.945
Sulphur
Nil
The chemical analysis of the nano material was
donebyCarbon-Hydrogen-Nitrogen-Sulphur(CHNS) test.
SAMPLE
PREPARATIONMETHODOLOGY

Fig. 1 Scanning electron microscopic image of fly
ash obtained by FEG-SEM

Test methodology: Modified proctor compaction
and Modified CBR test

Soil
Marine clay soil was obtained from the premises of
IIT, Bombay. Table 3 below represents the physicalproperties of soil.
Table 3 Physical properties of Soil
Properties
Specific gravity
Maximum dry density ( gms/cm3)
Optimum moisture content (OMC) %Gravel size
Sand size
Silt size
Clay size
Liquid Limit (%)
Plastic limit (%)
Shrinkage limit (%)
Plasticity Index
Group symbol
Coefficient of permeability ‘k’ cm/sec
Undrained shear strength (kPa)
Soaked CBR (%)

AND

Soil
2.57
1.63
22 %
-0.86 %
45.14 %
54 %
78
40.9
24.6
37.1
CH-MH
9.541 x 10-9
8.3
1.8

487

Modified proctor densities (simulating the heavy
compaction in field) were initially worked out as aprerequisite for conducting the modified CBR tests.
Oven dried fly ash was mixed to the soil inincreasing proportions by 5%,10%,15% and 20%
corresponding to 95%,90%,85% and 80% by dry
weight of soil. Further each proportion was treatedwith nano solution (nano material diluted in water)in four different dilution ratios of (1:150),(1:300),(1:600) and (1:800) by volume. The modified
proctor compaction tests were carried out on all
variations in accordance as in [28].
Before conducting the CBR tests, the treated soil
samples were allowed to interact with nano solution
for about 8 hours by covering in a polythene sheet.
The moulds were compacted to the respective
modified proctor densities and were left to air dry
for four days. The unsoaked CBR tests were
performed after 4 days of air drying. The moulds
were submerged in water for four days. The soaked
CBR tests were conducted after four days of

GEOMATE-Bangkok, Nov. 14-16, 2016

submergence in water. The samples tested formodified California bearing ratio were in accordancewith [29]. Fig.2 represents the modified CBR set up.

Table 5 Test variations considered in the program
Seriesno
SeriesI

Load cell

Dial gauge
SeriesII

Fig. 2 Modified CBR test set up.

Based on the results and optimization of modified
CBR tests the Atterberg’s limit tests were carried out
on proportions of Marine clay+ 10% fly ash withnano material dilution ratios of (1:150) and (1:300).
water) The nano solution treated sample wascovered in a polythene sheet and was left to interact
with the soil for 8 hours. The liquid limit, plastic
limit and shrinkage limit tests were conducted on all
the samples in accordance with [30][31].
Table 5 represents the proposed test variations
considered in the testing program.

95% Marine clay + 5% FA

SeriesIII

90% Marine clay + 10% FA
90%Marine clay + 10% FA with (1:150) nanodilution
90%Marine clay + 10%FA with (1:300) nanodilution
90% Marine clay + 10% FA with (1:600) nanodilution
90% Marine clay + 10% FA with (1:800) nanodilution

SeriesIV

85% Marine clay + 15% FA
85% Marine clay + 15% FA with (1:150) nanodilution
85% Marine clay + 15% FA with (1:300) nanodilution
85% Marine clay + 15% FA with (1:600) nanodilution
85% Marine clay + 15% FA with (1:800) nanodilution

Series
V

80% Marine clay + 20% FA
80% Marine clay + 20% FA with (1:150) nanodilution
80% Marine clay + 20% FA with (1:300) nanodilution
80%Marine clay +20% FA with (1:600) nanodilution
80% Marine clay + 20% FA with (1:800) nanodilution

RESULTS AND DISCUSSIONS
Modified California Bearing Ratio (CBR) tests
The Indian Roads Congress specifications forpavement design incorporate only the soaked CBRvalues of the subgrade material considering theworst conditions of submergence of the soil in water.
Therefore, only soaked CBR stress penetration
curves are being presented herewith. Figures 3 to 7
represent the Stress penetration curves for soaked
CBR samples.

Marine clay (neat soil)
Marine clay with (1:150) nano dilution
Marine clay with (1:300) nano dilution
Marine clay with (1:600) nano dilution
Marine clay with (1:800) nano dilution

95%Marine clay + 5% FA with (1:150) nanodilution
95%Marine clay + 5% FA with (1:300) nanodilution
95%Marine clay + 5% FA with (1:600) nanodilution
95%Marine clay + 5% FA with (1:800) nanodilution

Digital load output

Test methodology: Atterbergs limits

Sample details

488

GEOMATE-Bangkok, Nov. 14-16, 2016

800

600

marine clay with nanodilution(1:150)

700
Stress (kPa)Stress (kPa)

500

Marine clay with nanodilution (1:300)

400
300

Marine clay with nanodilution (1:600)

200

500

85%Marine
clay+15%Flyash withnano dilution(1:300)

400
300

85%Marine clay+15%Fly
ash with nanodilution(1:600)

100
0

0
0.0

2.5

5.0
7.5
Penetration (mm)

10.0

0

12.5

Fig. 3 Stress penetration curves for soaked CBRsamples of Series I.

800

500
400
300
200
100
2.5

5
7.5
Penetration (mm)

10

500

80%Marine clay+20% Fly
ash with nanodilution(1:300)

400
300

80% Marine clay+20% Fly
ash with nanodilution(1:600)

200
100

80% Marine clay + 20%
Fly ash with nano12.5
dilution(1:800)

0
0

12.5

2.5

5
7.5
Penetration (mm)10Fig. 7 Stress penetration curves for soaked CBRsamples of Series V.
Fig.8 depicts the graphical representation of
unsoaked and soaked CBR values considered in the
program from test series I to V.

90% Marine clay +10% Fly ash

1000
90% marine clay +10%
Fly ash with nanodilution(1:150)

800

80%Marine clay+20% Fly
ash with nanodilution(1:150)

600

Fig. 4 Stress penetration curves for soaked CBRsamples of Series II.
1200

80% Marine clay+20% Fly
ash

700

0
0

85%Marine clay+15% Fly
12.5
ash with nanodilution(1:800)

10

800

Stress (kPa)

600

7.5
5
Penetration (mm)900

95%marine clay+5% Fly
ash with nanodilution(1:150)
95% Marine clay+5%
Flyash with nano dilution(1:300)95% Marine clay +5% Fly
ash with nano dilution(1:600)95% Marine clay+5% Fly
ash with nanodilution(1:800)

700

2.5

Fig. 6 Stress penetration curves for soaked CBRsamples of Series IV.

95% Marine clay + 5%
Fly ash900

Stress (kPa)

85%Marine
clay+15%Flyash withnano dilution(1:150)

600

200

Marine clay with nanodilution(1:800)

100

90
Unsoaked
Unsoaked

80
Unsoaked

600

90% Marine clay +10% Fly ash with nanodilution(1:300)

400

0
0

2.5

5
7.5
Penetration (mm)10

without nanomaterial

Unsoaked
60

90%marine clay+10%
Flyash with nanodilution(1:600)

200

Unsoaked70

CBR (%)

Stress (kPa)

85% Marine clay+15%
Fly ash900

Marine clay (neat Soil)

700

50
with nanomaterial (1:150)

40
30

with nanomaterial (1:300)

20

90% Marine clay+10%
12.5 Fly ash with nanodilution(1:800)

SoakedSoakedSoaked10

SoakedSoaked0Marineclay

Fig. 5 Stress penetration curves for soaked CBRsamples of Series III.

95%
Marineclay +
5%
Flyash90%
MarineClay +
10%
Flyash85%
Marineclay +
15%
Flyash80%
Marineclay +
20%
Flyashwith nanomaterial (1:600)
with nanomaterial (1:800)

CBR test series I to V

Fig. 8 Graphical representation of unsoaked andsoaked CBR samples of Test Series I to V.

489

GEOMATE-Bangkok, Nov. 14-16, 2016

Table 6 represents the unsoaked and soaked CBRvalues of samples of Test Series I to V.

of soil in water. The CBR values were observed to
be increased with the addition of fly ash to the soil.
Soaked CBR was observed to be enhanced by
188.88% by adding 10% fly ash as compared to neatsoil. Further, both unsoaked and soaked CBR values
were observed to be decreased beyond 10% additionof fly ash to the soil. .

Table 6 Unsoaked and soaked CBR values of test
series I to V
Sr.
No

I

Sample details

II

Marine clay (neat soil)
Marine clay with (1:150)
nano dilution
Marine clay with (1:300)
nano dilution
Marine clay with (1:600)
nano dilution
Marine clay with (1:800)
nano dilution
95% Marine clay + 5% FA

III

95% Marine clay + 5% FA
with (1:150) nano dilution
95% Marine clay + 5% FA
with (1:300) nano dilution
95% Marine clay + 5% FA
with (1:600) nano dilution
95% Marine clay + 5% FA
with (1:800) nano dilution
90% Marine clay + 10% FA

Unsoaked
CBR
(%)
50.14
57.47

Soaked
CBR
(%)

59.01

5.39

56.12

3.47

52.07

2.95

54.89

3.08

62.67

6.42

72.12

7.13

69.42

4.5

61.52

2.7

58.24

5.2

On further treating the samples with nanomaterial in four different dilutions of (1:150),(1:300), (1:600) and (1:800) the CBR values wereobserved to be significantly improved. With 10% fly
ash and nano dilution ratio of (1:150) the soaked
CBR value was observed to be enhanced by
381.66% and with (1:300) nano dilution ratio thesoaked CBR was enhanced by 478.33% as comparedto the neat soil. In all the samples of tests series I to
V it was observed that the dilution ratio played avital role in improvement of CBR values.

1.8
4.37

The significant increase in CBR values is due tothe formation of strong bonds when the soils comein contact with organo silane nano material. The
stronger long chain bonds are formed on hydrolysis.
After breakage of these bonds further affinity ofsilica towards the oxygen forms stronger bonds
which will become parting a good hydrophobicity to
the soil surface. Since the Class ‘F’ fly ash being
used in this research is very rich in silica (62.8%),
more and more silica becomes available for theinteraction with nano material thereby forming morestrong bonds within the soil mass.

90% Marine clay + 10% FA 75.59
8.67
with (1:150) nano dilution
90% Marine clay + 10% FA 82.52
10.41
with (1:300) nano dilution
90% Marine clay + 10% FA 73.4
7.32
with (1:600) nano dilution
90% Marine clay + 10% FA 63.25
6.55
with (1:800) nano dilution
IV 85% Marine clay + 15% FA 55.42
3.1
85% Marine clay + 15% FA 70.39
7.52
with (1:150) nano dilution
85% Marine clay + 15% FA 74.24
7.9
with (1:300) nano dilution
85% Marine clay + 15% FA 62.87
6.17
with (1:600) nano dilution
85% Marine clay + 15% FA 60.36
5.0
with (1:800) nano dilution
V
80% Marine clay + 20% FA 51.1
2.12
80% Marine clay + 20% FA 68.27
6.17
with (1:150) nano dilution
80% Marine clay + 20% FA 67.11
6.92
with (1:300) nano dilution
80%Marine clay +20% FA
61.13
3.47
with (1:600) nano dilution
80% Marine clay + 20% FA 59.52
2.18
with (1:800) nano dilution
The soaked CBR values are very lesser as comparedto the unsoaked values on account of submergence

Atterberg’s limit tests
Table 7 summarizes the Atterberg’s limit results. LL,
PL and SL values were observed to be decreased ontreating the soil with 10% fly ash. Further on treating
the sample (marine clay + 10% fly ash) with nanomaterial dilution ratio of and (1:300) the LL, PL, SL
and PI values significantly decreased by 40.87%,
41%, 37.39% and 49.05% respectively.
This
indicated a substantial decrease in the plasticity
characteristics of soil thereby transforming thehighly plastic soil to medium plastic soil as per
Indian standard soil classification [32].

490GEOMATE-Bangkok, Nov. 14-16, 2016stronger long chain bonds, thus imparting goodhydrophobicity to the soil and improving the CBRproperties.
6. The research is thus an attempt to stabilize the soil
together with fly ash and nano material which holdsa very promising future for pavement infrastructural
applications.

Table 7 Atterberg’s limits test results
SampleLL
PL
SL
PI
description(%) (%)(%)(%)
Marine clay
78
40.9 24.6 37.1
(neat soil)
Marine clay withnano dilution
58
30.6 22.7 27.4
(1:150)Marine clay withnano dilution
51
28.7 18.6 22.3
(1:300)Marine clay +
62
31
23.2
31
10% Fly ash
Marine clay +10% Fly ash with
46
28
18.61
18
nano dilution(1:150)Marine clay +10% Fly ash with43
24.1 15.4 18.9
nano dilution(1:300)The improvement in plasticity of soils can be
attributed to the formation of strong bonds. They
impart hydrophobicity to the soil and make it morewater repelling. The new bonds so formed neutralize
the net negative charge on the silt and clay particles[25], thereby reducing the plasticity of soil making it
more compatible to be handled on site.

ABBREVIATIONS
The following symbols are used in this paper.
CBR = California Bearing Ratio
FA = Fly ash
LL = Liquid limit
PL = Plastic limit
SL = Shrinkage limit
PI = Plasticity Index
ACKNOWLEDGEMENT
The nano material used in the research wassponsored by Zydex Industries, Vadodara, Gujarat,
India. The authors acknowledge the support andcooperation given by Zydex Industries during thisresearch work.
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properties of texturally variant soils”,
Bioresource Technology 75(2000) 91-93.
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[5] Bhuvaneshwari, S., Robinson, R.G. and Gandhi,
S.R.(2005), “ Stabilization of expansive soils
using Fly ash”, Fly ash Utilization Programme(FAUP), TIFAC, DST, New Delhi,India.
[6] Consoli, N.C., Prieto, P.D.M., Carraro.,J.A.H,
Heineck, K.S,( 2001), “ behaviour of compactedsoil- fly ash- carbide lime mixtures”, Journal of
Geotechnical and Geo environmental
Engineering, 2001.127: 774-782.
[7] Kaniraj,S.R. and Havanagi, V.G (2001),“
Behavior of cement stabilized fiber reinforced
fly ash –soil mixtures”, Journal of Geotechnicaland Geo-environmental Engineering, 2001.127 :pp 574-584.

Suitability as a good subgrade material
The soil when treated with fly ash and nano material
is seen to improve its CBR properties to a great
extent thereby satisfying the subgrade criteria of
CBR of 8% for roads having traffic of 450
commercial vehicles per day pertaining to the Indian
sub continent as in [33]
CONCLUSIONS
1. Class ‘F’ Fly ash is found suitable for stabilizationof soft soil. Soaked CBR values were found to be
increased from 1.8 to 5.2 which was substantially
improved by treating the soil with 10% fly ash.
2. Maximum improvements were observed in soil
treated with 10% fly ash and nano material withdilution ratios of (1:150) and (1:300).
3. With 10% fly ash and dilution ratio of (1:300) the
soaked CBR value was observed to be improved to10.41 which indicated a good improvement,
compared to the unreinforced soil.
4. LL, PL, SL and PI values significantly decreaseby 40.87%, 41%, 37.39% and 49.05% respectively
on treating the soil with 10% fly ash and nanodilution ratio of (1:300) thus transforming the highly
plastic soil to medium plastic soil.
5. Class ‘F’ fly ash adopted in the study was foundto be highly siliceous which helps in formation of491

GEOMATE-Bangkok, Nov. 14-16, 2016[8] Arvind kumar, Walia, B.S. and Bajaj, A.(2007),“Influence of fly ash, lime and polyester fibers oncompaction and strength properties of expansive
soils”, Journal of Materials in Civil Engineering,
2007.19: 242-248.
[9] Shafique, S. B., Edil, T.B., Benson,C.H. andSenol, A.(2004), “ Incorporating a fly ash
stabilized layer in to pavement design”,
Proceedings for the Institution of Civil
Engineers, Geotechnical Engineering, 157 Issue
GE4.
[10] Dutta, R.K. and Sarda, V.K..(2006), “CBR
behavior of Waste plastic strip-reinforced stone
dust/ fly ash overlying saturated clay,Turkish
Journal of Engineering and Environmental
Science, 31(2007), 171-182.
[11] Lin, D. F., Lin, K. L., Hung, M. J. and Luo, H.
L.(2007).“Sludge ash / hydrated lime on thegeotechnical properties of soft soil” , Journal ofHazardous Materials, Vol. 145(1-2), 25 June
2007, pp. 58-64.
[12] Saha, S. and Pal, S.K.(2013),” Influence of Fly
ash on Unconfined compressive strength of soil
and fly ash layers placed successively”,
Electronic Journal of GeotechnicalEngineering,Vol.18[2013], Bund.H, pp:1593–
1602.
[13] Nadaf, M.B. and Mandal, J. N (2013), “Experimental studies and Analysis for Basic
Characterization of Fly ash”, Proceedings of
4th Global Engineering, Science andTechnology Conference, 27-28 December,
2013, BIAM Foundation, Dhaka, Bangladesh,
ISBN: 978-1-922069-43-6.
[14] Ram Rathan lal, B. and Mandal, J. N. (2013),”
Study of Cellular reinforced fly ash under
triaxial loading conditions”, International
Journal Of Geotechnical Engineering, 2013;
7(1): pp: 91–104.
[15] Kulkarni,P.P. and Mandal, J.N.(2016),” CBR
behaviour of soil treated with Class ‘F’ Flyash”,
STM Journal of Geotechnical Engineering,
ISSN:2394-1987,Volume 3,Issue 1, pp: 33-43.
[16] Yunus,N.Z.M, Marto,A., Pakir,F.,Kasran,K.,
Jamal,M.A.K, Jusoh,S.N. and Abdullah,N, “
Performance of lime treated marine clay on
strength and compressibility characteristics”,
International Journal of Geomate, June,
2015,Vol.8, No.2 (Sl.No.16), pp. 1232-1238.
[17] Patel,S. and Shahu,J.T, “ Engineering properties
of Black cotton soil-dolime mix for its use as
subbase material in pavements”, International
Journal of Geomate, March, 2015,Vol.8, No.1
(Sl.No.15), pp. 1159-1166.
[18] Li, Gengying.(2004), “ Properties of highvolume fly ash concrete incorporating nanoSiO 2 ,” Journal of Cement and Concrete
Research, 34(2004) 1043-1049.
[19] Luo, H.L., Lin, D.F. and Kuo, W.T. (2004),“The effects of nano materials on the behaviourof sludge mortar specimens”, Water Science andtechnology, 0273-1223.
[20] Chen ,L and Lin,D.F.(2009), “Applications ofsewage sludge ash and nano SiO 2 to
manufacture tile as construction material”,
Construction and Building Materials 23(2009)
3312-3320.
[21] Luo, H.L., Hsiao, D. H., Lin, D. F. and Lin, C.
K. (2012). “Cohesive Soil Stabilized Using
Sewage Sludge Ash/Cement and Nano
Aluminum Oxide,” International Journal ofTransportation Science and Technology, Vol.
1(1), pp. 83-100.
[22] Doshi, R., Braida, W., Christodoulatos, C.,
Wazne, M. and Connor, G.(2008), “ Nanoaluminium : Transport through sand columns
communities”, Journal of Environmental
research 106(2008) 296-303.
[23] Nazari, A. and Riahi, S. (2011). “The effects ofzinc dioxide nano particles on flexural strengthof self-compacting concrete , Composites: Part
B, Vol. 42, pp. 167-175.
[24] Kawashima, H., Hou, P., Corr, D.J. and Shah,
S.P.(2012), “ Modification of cement basedmaterials with nanoparticles”, Cement andConcrete composites 36(2013) 8-15.
[25] Ugwu, O.O., Arop, J.B., Nwoji, C.U. andOsadebe, N.N.(2013),“Nano technology as apreventive engineering solution to highway
infrastructure failures”, Journal of Construction
Engineering and Management, 2013. 139:987993.
[26] Moradi, G. and Seyedi, S. (2015), “Effect Of
Sampling Method on Strength of Stabilized
Silty sands with colloidal Nano Silica”, Journalof Civil Engineering Research 2015, 5(6): 129135.
[27] ASTM C618-08a.2008. Standard specification
for coal fly ash and raw or calcined natural
pozzolan for use in concrete.
[28] Indian Standard IS 2720 (Part 8)-1983, Methodsof test for soils, Part 8: Determination of watercontent- dry density relation using heavy
compaction.
[29] Indian Standard IS 2720(Part16) -1987,
Methods of test for soils, Part 16: Laboratory
determination of CBR.
[30] Indian Standard IS 2720 (Part5)- 1985,
Methods of test for soils, Part 5: Determinationof liquid and plastic limit.
[31] Indian Standard IS 2720(Part6)- 1972, Methodsof test for soils, Part 6: Determination of
Shrinkage factors.
[32] Indian Standard IS 1498 -1970, Classificationand identification of soils for generalengineering purposes.
[33] IRC:37-2012 (Tentative guidelines for design offlexible pavements).
492

GEOMATE- Bangkok, Nov. 14-16, 2016

Environment493

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

USING FLOATING WETLAND TREATMENT SYSTEMS TO REDUCESTORMWATER POLLUTION FROM URBAN DEVELOPMENTS
Peter Schwammberger1, Chris Walker2 and Terry Lucke1
Stormwater Research Group, University of the Sunshine Coast, Australia
2
Covey Associates Pty Ltd, Sunshine Coast, Australia

1

ABSTRACT
Floating treatment wetland (FTW) systems are an innovative stormwater treatment technology currently being
trialled in Australia. FTWs provide support for selected plant species to remove pollutants from stormwater
discharged into a storage basin. The plant roots provide large surface areas for biofilm growth, which serves totrap suspended particles and enable the biological uptake of nutrients. FTWs can be installed at the start of theconstruction phase and can therefore start treating construction runoff almost immediately. FTWs have thepotential to provide a full range of stormwater runoff treatment (e.g. sediment and nutrient removal) from theconstruction phase onwards. A 101 m2 FTWs has been installed within a greenfield development site on theSunshine Coast, in Australia. The two-year research study investigated the pollution removal performance of the
FTW for two different locations, one with low and one with moderate influent pollutant concentrations. This paper
presents the research methodology used, and the initial study results of the treatment efficiency of FTWs.
Keywords: urban stormwater runoff, floating treatment wetlands, stormwater pollution, stormwater treatment
INTRODUCTIONAs FTWs are buoyant, they can be installed inmost water bodies without significant earthworks,
thereby offeringdevelopers
aneffective,
environmentally sustainable and cost-effective water
treatment solution [9]. FTWs also offer terrestrial andaquatic habitats for wildlife which allows for greater
ecological diversity [3].
Plant growth, establishment and survival inconventional stormwater treatment devices (e.g.
constructed wetlands) are often affected by high flow
velocities, the duration and depth of inundation, andthe frequency of flooding or drought [10].
Consequently, the wetland area may need to berelatively large to buffer against these extreme waterlevel fluctuations. Alternatively, the high flows may
be designed to bypass the wetland all together
resulting in a significant portion of stormwater runoff
being untreated [11].
The buoyancy of FTWs generally enables them totolerate major fluctuations in water depth. This makes
FTWs a reliable stormwater treatment device that
protects plant health during extreme rainfall events.
Furthermore, the extensive surface area of the root
network can provide significantly greater pollution
removal rates per unit area compared to constructedwetlands and other stormwater treatment systems.
Several mesocosm studies have been conducted toinvestigate the treatment efficiency of different plant
species planted in FTWs. These studies have shownthat FTW are effective at removing pollutants fromboth wastewater [12-15] and stormwater [7,16,17].
The number of field studies undertaken on FTWs islimited, particularly in the application of FTWs tostormwater treatment [3,18,19]. Smaller-scale field

Natural floating wetland ecosystems exist in
many parts of the world, ranging from large floating
marshes covering thousands of hectares in Louisiana[1] to smaller floating mires in the Netherlands [2].
Floating Treatment Wetlands (FTWs) are artificial
systems that mimic the water treatment processes thattake place in natural floating wetland islands [3].
FTWs are comprised of a floating structure plantedwith emergent macrophytes where the roots grow into
the water column [4]. The roots provide a large
surface area (Figure 1) for biofilm attachment. The
biofilm adsorbs pollutants [5,6] while physicallyfiltering and trapping suspended solids [7].

Fig. 1Figure showing root hairs and vegetation
above the floating mat [8].
494GEOMATE- Bangkok, Nov. 14-16, 2016

studies in the USA [19] and New Zealand [4] haveshown FTWs to be effective in removing a variety ofpollutants from stormwater runoff from highways.
To date, no previous studies have been undertaken
in an urban residential setting, where there can behigher pollutant loads such as sediment, nutrients, andheavy metals. These urban stormwater pollutants canbe readily transported to receiving waters due to anincrease in impervious surfaces such as roads,
sidewalks, driveways, parking lots and rooftops, aswell as more efficient drainage pathways [6]. This
paper describes the results of a field study conducted
to assess the ability of FTWs to treat urban
stormwater runoff from an existing urban
development in Southeast Queensland (SEQ), in
Australia.

The study involved installing a new FTW in theexisting lake shown in Figure 2. The FTW had a total
area of 101 m2 which was equivalent toapproximately 1.7% of the development area. The
FTW was comprised of 11 FTW modules, each of 9
m2 in area. The FTWs (www.spel.com.au) used inthis study were composed of a 200 mm thick,recycled plastic fibre mat, injected with marine grade
foam to provide buoyancy (Figure 3). Each mat had
40 pre-drilled holes to hold the plants. The holes were
100 mm in diameter and 150 mm deep. The mats werecovered with coir matting and planted with tube
stocks of Carex appressa at a density of two plants
per hole (Figure 3).
To direct the stormwater inflow into the lake andthrough the FTWs to limit potential short-circuitingof the FTW (i.e. bypassing flows) experienced inother studies [11,18], impermeable polyvinyl
chloride (PVC) curtains were installed along the sidesof the FTWs that fanned out towards the lake banks(Figure 2). These ensured all inflows from Inlets 1
and 2 were treated by the FTWs.

MATERIAL AND METHODS
Study Site
This study was conducted in an existing lake
within a development under construction on Bribie
Island in Queensland, Australia. The entire
development site has an area of 42.3 ha, withconstruction having commenced in 2014. At the startof the research project in September 2014,stormwater from a 10ha existing residential
catchment (external to the new development)
discharged through two existing inlets into the study
lake (Inlets 1 and 2 in Figure 2).

Fig 3

Fig 2

Two of the FTW modules planted with
Carex appressa.

Water quality analysis results from the firstsampling period (from September to February 2015)
showed that the influent pollutant concentrations
were very low. In late February 2015, the study site
was impacted by a tropical cyclone (Marcia - TCM)
which caused some minor damage to the FTW. In
particular, some of the anchoring cable supports weredamaged which caused the FTW to shift its location.
Portions of the PVC curtain were also damaged.
As the FTW needed realignment and repair after
TCM, it was decided to use this opportunity to
relocate the FTW, and to change the baffle
configuration to enable the treatment of stormwater
inflows from a different part of the catchment which
was thought to have potentially higher pollution
concentrations. This resulted in relocating the FTW
to the new position shown in Figure 4. The newlocation meant the FTW received stormwater (Inlet 3)from a 5.3 ha existing residential area, as well as
stormwater from a new 2.2 ha area of the Bribie Lakes
development (Total treatment area = 7.5 ha). As the
new 2.2 ha area was under construction, it was

Initial location of the FTW and automatic
samplers (AS).
495

GEOMATE- Bangkok, Nov. 14-16, 2016

expected to have much higher influent pollutant
concentrations. Prior to the reconfiguration of the
FTW, the discharge from Inlet 3 bypassed the FTW.

Fig 4

Storm Event Sampling and Analysis
In order to standardize the sampling methods
and procedures to capture qualifying storm events, asampling protocol (Table 1) was developed based on
the protocol methods prescribed by the United States
Environmental Protection Agency’s (US EPA)
Stormwater BMP Monitoring Manual [20] andAuckland Regional Council’s Proprietary Device
Evaluation Protocol for Stormwater Quality
Treatment Devices [3,21].
Automatic water samplers (ISCO GLS www.isco.com) were installed upstream (AS1) anddownstream (AS2) of the FTW (Figures 2 and 4) tocollect water samples to be analysed for water quality
parameters (Table 1) by a NATA (National
Association of Testing Authorities) accredited
laboratory. The samples were analysed for totalsuspended solids (TSS), total nitrogen (TN) and total
phosphorus (TP) which are the three main stormwaterpollutants of concern in Australia. The auto-samplers
were triggered by a combination of signals from a rain
gauge and flow meter that were installed on site. The
auto-samplers were programmed to collect 200 mL
flow-weighted aliquots each time a cumulative
volume of 15 kL registered at the flowmeter. This was
equivalent to the runoff from a rainfall depth of 0.2
mm over the study catchment area. The 200 ml
aliquots were composited into a nine litre glass bottle
and sent to the laboratory for analysis.

New location of the FTW and automatic
samplers (AS) after TCM

Table 1

Sampling Protocol and Analysis Details

Parameter
Minimum rainfall depth
Rainfall monitoring
Minimum storm duration
Minimum antecedent dry period
Minimum hydrograph sampling
Minimum number of sample aliquots
Sampling method
Sampler location
Total Suspended Solids (TSS)& Total Dissolved Solids (TDS)
Total Nitrogen & TKN
Ammonia N
NOx
Total Phosphorous & Orthophosphate
Particle Size Distribution (PSD)

Details
2mm in 30 min
Pluviometer
15 min
6 hours
First 60% of hydrograph
Minimum 8 influent and 8 effluent subsamples per event
ISCO GLS Auto-samplers, flow-weighted in 15 kL intervals
1.3 m upstream and 3.3 m downstream of FTWAPHA (2005) 2540 C & D
APHA (2005) 4500 N
APHA (2005) 4500 NH3
APHA (2005) 4500 NO3
APHA (2005) 4500 P
Laser Diffraction (Malvern Mastersizer 3000)

496

GEOMATE- Bangkok, Nov. 14-16, 2016

The water quality analysis results were used toestimate an Event Mean Concentration (EMC) for
each storm event using Equation 1:

 VC
EMC 
 V

from 0.41 to 0.91 mg/L for TN, and from 0.005 to
0.078 mg/L for TP. These concentrations were well
below the ‘typical’ Australian average urban
pollutant loads expected in SEQ which are 151 mg/L
for TSS, 1.82 mg/L for TN and 0.34 mg/L for TP[22].
For events prior to TCM, the average pollutant
removal efficiency ratios (ER) for TSS, TN and TP
were -5%, 16% and 41%, respectively (Table 2).
These were below the minimum pollutant removal
ERs recommended for urban developments in SEQ of
80%, 60% and 45%, for TSS, TN and TP respectively[23]. The variation in the results between theindividual storm events was substantial and pollution
removal ER ranged from between -88% and 80% for
TSS, -6% and 47% for TN and between 0% and 59%
for TP. However, these results needs to be considered
carefully, and within the correct context.
For, example, the 41% ER results for TP removal(Table 2) needs to be considered with care as there
was no net removal of TP for three of the five eventsand these were below detection limits. The highvariability of the results was a result of the low
pollutant concentrations generated from the existing
residential development.
Furthermore, it was thought that the 70 m lengthof open water in the lake in front of the FTW wasprobably acting as a type of pre-treatment system forthe inflowing stormwater leading to sedimentation,and portions of TSS and particle bound pollutants notreaching the FTW.
Once the location of the FTW was changed after
TCM to treat the stormwater from Inlet 3 (Figure 4)
water quality sampling results show much higher inlet
pollution concentrations (Table 3).

ni 1 ini 1i

Eq. (1)iwhere:
Vi = Volume of flow during period i;
Ci = Concentration associated with period i; andn = total number of aliquots collected during event.
The water quality analysis results were also used
to assess the overall system performance by
calculating the pollution removal efficiency ratio(ER) using Equation 2:

ER  1 

Mean EMC Out
Mean EMC In

Eq. (2)where:
EMCin = Event Mean Concentration at AS1 and
EMCout = Event Mean Concentration at AS2.

RESULTS AND DISCUSSION
Sampling commenced in September 2014 andcontinued through to February 2015 when the studysite was affected by TCM. During this time, fivequalifying storm events were captured. As shown in
Table 2, results prior to TCM showed low
concentrations of TSS, TN and TP. The
concentrations ranged from 4 to 10 mg/L for TSS,
Table 2

Pollution removal efficiency prior to TCM

ParameterDate
07/09/14
23/09/14
25/09/14
07/11/14
20/11/14
Mean Conc.
Efficiency Ratio

TSS(mg/L)
In out
10 14
5
4
4
4
10
2
8
15
7.4 7.8-5%TN(mg/L)
Inout
0.51 0.53
0.41 0.44
0.41 0.43
0.65 0.56
0.97 0.52
0.59 0.49
16%TP(mg/L)
Inout
0.005 0.005
0.005 0.005
0.005 0.005
0.052 0.038
0.078 0.032
0.029 0.017
41%

The inlet pollutant concentrations ranged from 11
to 414 mg/L for TSS, from 0.6 to 3.2 mg/L for TN
and from 0.03 to 0.28 mg/L for TP. These results are
more in-line with the expected pollutant
concentration ranges for stormwater runoff fromurban developments in SEQ. [22].

Peak Flow(L/s)

Inflow Volume
(m3)

54
n/a
34
41
34

890.6
n/a
433.1
169.9
367.8

The average pollutant removal efficiency ratios
for TSS, TN and TP were 80%, 17% and 52%,
respectively. Compared to the initial setup where TSS
concentrations often increased when passing throughthe FTW, the results after TCM show much better
TSS removal for higher influent concentrations, aswell as for higher flow rates (Table 3).
497

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 3
ParameterDate
28/09/15
23/10/15
07/11/15
15/11/15
29/11/15
30/01/16
01/02/16
06/02/16
13/02/16
06/03/16
Mean Conc.
Efficiency ratio

Pollution removal efficiency post-TCM
TSS(mg/L)
In
out
323
51
11
4
414
24
26
16
270
28
50
26
19
36
19
24
37
19
56
15
122.5 24.3
80%

TN
(mg/L)
Inout
1.00 0.25
0.70 0.30
3.20 0.70
1.10 0.70
2.20 1.30
1.10 2.20
0.80 1.60
0.60 0.80
1.40 2.10
1.20 1.10
1.33 1.1117%TP(mg/L)
In
out
0.280
0.1
0.030 0.02
0.280 0.03
0.050 0.05
0.140 0.02
0.040 0.04
0.040 0.07
0.050 0.03
0.060 0.04
0.100 0.11
0.107 0.051
52%

It is difficult to accurately assess the efficiency oftreatment systems when pollution concentrations areclose to detection limits, as even a small change inconcentration results in a significant change in the
average results. The relocation of the FTW after
TCM, and the subsequent increase in pollutant
loading, resulted in a higher pollutant removal ratio
for all pollutants monitored in the study.
The FTW showed a high variability in treatment
performance for stormwater runoff with very low
pollutant concentrations. However, when thepollutant loads were closer to those typically expected
for stormwater in SEQ, the ER was substantiallyhigher. The results of this study demonstrate that the
sampling location, and the influent pollutant loads are
extremely important. The study showed that thesevariables can significantly influence the results ofperformance and efficacy measurements of FTW
systems.
To further investigate the stormwater treatment
performance of FTWs, a new four-year researchstudy has recently commenced in a new development
site on the Sunshine Coast in Queensland, Australia.
The new study will be conducted on a 2,100 m2 FTW
treating stormwater runoff from a new 95 haresidential development. The stormwater runoff
water quality will be characterised and assessed
during both the construction and the operational
phases of the development. The ability of FTWs to
improve stormwater quality, and to manage urban
lake health will be evaluated throughout the four-year
study.
The study results will be of significant interest forthe developers and local government authorities andcould potentially influence stormwater management
practices, both in Australia, and internationally.

Peak Flow(L/s)

Inflow Volume
(m3)

n/a
100
1,340
2,335
1,160
24
93
126
320
151

n/a
279.3
511.9
531.0
361.0
234.8
281.0
611.6
611.4
148.6

CONCLUSION
This study initially investigated the pollutant
removal performance of a FTW receiving stormwaterrunoff from a 10 ha residential site. However,stormwater pollutant concentrations from the sitewere found to be far below typical concentrations forurban stormwater runoff in Australia and this resulted
in low pollutant removal performance results for the
FTW.
After approximately six months, the FTW was
moved to a new location (catchment area = 7.5 ha)with higher influent pollutant concentrations thatwere more in-line with typically expected values forurban catchments in Australia. The pollutant removal
performance of the FTW in the new location
increased significantly and the average pollutant
removal efficiency ratios were found to be 80%, 17%
and 52%, respectively for TSS, TN and TP. These
results were much closer to the recommended
removal rates in Australia.
The study results demonstrate that sampling
location, and influent pollutant loads are extremely
important and that these variables can significantly
influence the results of performance and efficacy
measurements of FTW systems.
The study has demonstrated that FTW are a viableoption for urban stormwater treatment that havenumerous advantages compared to traditional
systems. These include reduced land requirements,
resilience to extreme water depth and volume
changes, as well as potentially enhancing habitat,
recreational, and aesthetic values within the urban
landscape. It is anticipated that the study results could
significantly influence the stormwater management
in Australia and the rest of the world.

498

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[4] Borne KE, “Floating treatment wetland
influences on the fate and removal performanceof phosphorus in stormwater retention ponds”,
Ecolog. Engineering, Vol. 69, 2014, pp. 76-82.
[5] Headley TR, Tanner CC, “Constructed Wetlands
with Floating Emergent Macrophytes: An
Innovative Stormwater Treatment Technology”,
Critical Reviews in Environmental Science andTechnology Vol. 42 (21), 2012, pp. 2261–2310.
[6] Lynch J, Fox LJ, Owen JS, Sample DJ,“Evaluation of commercial floating treatment
wetland technologies for nutrient remediation ofstormwater”, Ecolog. Engineering, Vol. 75, 2015,
pp. 61–69.
[7] Headley TR, Tanner CC, “Floating wetlands forstormwater treatment: Removal of copper, zincand fine particulates”, Auckland Regional
Council Technical Report, 2008, New Zealand.
[8] Royal Botanic Gardens Melbourne (RBGM)(2016), Plant Collections, Lake System Wetlands,
website,accessedon
15-5-2016at:https://www.rbg.vic.gov.au/visitmelbourne/attractions/plant-collections/lakesystem-wetlands[9] Hartshorn N, Marimon Z, Xuan Z, Cormier J,
Chang NB, Wanielista MP, “Complex
interactions among nutrients, chlorophyll-a, andmicrocystins in three stormwater wet detention
basins with floating treatment wetlands”,
Chemosphere, Vol. 144, 2016, pp. 408–419.
[10] Chang NB, Islam MK, Wanielista MP, “Floating
wetland mesocosm assessment of nutrient
removal to reduce ecotoxicity in stormwater
ponds”, Int. J. Environ. Sci. Technol. Vol. 9 (3),2012, pp. 453–462.
[11] Headley TR, Tanner CC, “Application of
floating wetlands for enhanced stormwater
treatment: A review”, NIWA Client Report, 2006,
Auckland Regional Council, New Zealand.
[12] Ash R, Truong P, “The use of Vetiver grass
wetlands for sewerage treatment in Australia”,
3rd International Conference on Vetiver, 2003,
Guangzhou, China.
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PROPOSITION OF A DETERIORATION PREDICTION MODEL
FOR MAINTENANCE OF RUNWAY PAVEMENT
Hiroto Akimoto1, Naoaki Suemasa1, Kazuya Itoho1, Tsuyoshi Tanaka1
1
Urban and Civil Engineering, Tokyo City University, Tokyo, Japan;
ABSTRACT
Many airports have been built in Japan during the high economic growth period under the policy of “One airport
in each prefecture”. As a result of this policy, 102 airports are dotted around Japan currently. The number ofdomestic airline users tends to increase drastically since the air transportation has become more convenient andaffordable because of the improvement of long-range flight and the new service of LCC. On the other hand,according as the demand for planes has increased, the increase of the taking-off and landing number greatly
influences the deterioration of runway pavement. For that reason, it is essential to provide runways with efficientmaintenance to prevent from air accidents. However, a deterioration prediction technique for runways has not yet
been established at the moment. Therefore, a simulation model to estimate deterioration degree of runway
pavement was proposed in this study. Three elements such as a crack, a rutting and flatness are closely related to
the runway deterioration and a deterioration prediction technique was examined focusing on a crack percentage
and a rutting in this study. The future progress of deterioration degree was predicted through studying the actual
values obtained from the runway pavement in Kumamoto Airport. As a result, a deterioration prediction method
was confirmed by using the equation for determining the degree of the asphalt pavement fatigue. In addition, theessential elements intimately related to degradation were able to be extracted through the analysis of data from
Kumamoto Airport
Keywords: Run way，Asphalt pavements，Deterioration rate，Crack percentage
INTRODUCTION
It has been more than 50 years since many facilities
and buildings were constructed during the rapid
economic growth period in Japan. Nowadays, theburden of maintenance for such structures isgradually increasing. Maintenance is regarded asimportant especially for the airport facilities which
scatter in 102 places in the whole of Japan. Fig. 1
shows the change in the number of airline users inJapan [1]. As seen in the figure, the number of airline
users and the demand for plane increase. The
transportation using planes has increased due to theimprovement of long range flight ability and theactivation of airport industry by involving newservices such as low cost carrier. Accordingly, thenumber of planes taking off and landing has increased,
too. On the other hand, the increase in runway
consumption greatly influences the deterioration of
pavement. While the number of the check points formaintenance has increased, it has become moredifficult for some airports such as Haneda Airport tofind enough time to be checked because takeoffs andlandings of planes are continuously carried out for 24hours. It is, therefore, imperative to efficiently
perform the maintenance of runway. Since a
deterioration prediction method for runway has notbeen established at present, a degradation prediction

model for runway pavement was suggested in thisstudy aiming at improving the efficiency of its
maintenance.
A prediction model for cracks on runway was madein the study. The factors such as temperature, runningspeed and a load of plane were used for constituting
the model, which tend to greatly influence the runway
crazing. Data was based on the degradation progress
degree of the crack percentage of the runway actually
measured in Kumamoto Airport. The results from themodel were compared with the check data anddiscussed.

Fig. 1 Change in the number ofairline users in Japan
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CHANGE OF CRACK PERCENTAGE WITH
TIME LAPSE AND CURRENT DEGRADATIONSTATE IN KUMAMOTO AIRPORT

It can be confirmed from the elevated crack
percentage in the figure that degradation rate has
increased with time elapsed. The factor of cracking
on asphalt pavements is assumed to be mainly thefatigue by traffic load since the load on runway
pavement is extremely greater than that on generalroad. Especially as for oil asphalt pavement, cracks
hardly occur in the initial state, however, once thepavement has become hard after losing oil, cracks
tend to occur more easily.
It is thought to be possible to predict themechanism of runway deterioration by investigating
the distribution of crack percentage with time elapsed.

In the case of runway where heavy planes such as
jumbo jets take off and land, a section of the runway
with 30m in length and 21m in width is defined as one
unit and check and repair of runway are carried out
by every unit. Kumamoto Airport adopts this unit
system because the length of the runway is 3, 000m
and jumbo jets goes into service on the runway. Fig.
2 shows the change of crack percentage measured
from 1984 through 2008 at Kumamoto Airport.

CRAZING DEGRADATION SPEED
DEGRADATION PREDICTIONAND

The progress degree of deterioration in each unit
can be estimated by the slope of the degradation
increase from measured values of crack percentage.
For Kumamoto Airport, the crack percentage
increment after the whole repair performed in 1994
was referred as the measured values. Since the changeof the crack percentage with time increased with awide variety of patterns as shown in Fig. 3, it wasdifficult to clearly evaluate the vulnerability to cracks
by each unit. Therefore, in order to facilitate thecomparison of the damage by each unit, it was neededthat the progress degree of deterioration which nonlinearly increases should be linearized. So, a powerfunction was introduced to linearize the relationshipbetween the progress degree of deterioration andelapsed month, as shown in the following equation.
The progress degree of deterioration in crack
percentage and the slope of increasing crack
percentage were defined as a crack index, Icr, and acrack deterioration speed, vcr, respectively.

Fig. 2 Relations of year and crack percentage
The red dotted line and the blue dotted line in the
figure indicate the damage levels determined by arepair evaluation standard for runway. The red dottedline is the C evaluation level, meaning that prompt
repair is required, while the blue dotted line is the Aevaluation level, meaning that prompt repair is not yet
required. Because repair was given at Kumamoto
Airport in the early 1990s, the crack percentage
decreased once at the time, however, sudden increase
was seen after the year and reached the C evaluation
at some units in 2008. Fig. 3 shows the relationshipbetween elapsed month and crack percentage startingfrom 1994 when repair was performed.

R1   I cr

(1)

I cr  vcr  T(2)where R: crack percentage (%), μ: the power
number, Icr: the crack index, vcr: the crack
deterioration speed, T: the elapsed month after repair.
As shown in the equation (2), the crack index “Icr” isdefined as the value of the crack percentage, R, raised
to the power of Icr should be linearly proportional
to the elapsed month after repair, T, and its
proportional coefficient is defined as the crack
deterioration speed “vcr”. From the data of the power
best fitted in each unit, the frequency distribution ofFig. 3 Relations of elapsed time and crack percentage

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GEOMATE- Bangkok, Nov. 14-16, 2016

the power value was made and approximated with thenormal distribution curve. Finally, the power number
used in the following discussion was chosen from themost frequent value of the curve. The frequency
distribution and the normal distribution of power
value for the crack percentage are shown in Fig. 4.

Fig. 6 Runway distribution of crack deterioration
speed.
It was cleared from the distribution that thedegradation speed was high in the vicinity of the unit
20 and low in the vicinity of the unit 70. Fig. 7 shows
the crack indexes of both the most and the least
vulnerable units to deterioration.

Fig. 4 A frequency distribution and normal
distribution of power value in crack
percentage
From this result, 2.41 was obtained as its power
number. Then, the crack index “Icr” for each unit wascalculated using the power function as shown inequation (2), and the crack deterioration speed “vcr”
was determined from the slope of crack index Icr toelapsed time. In order to compare among thedegradation speeds, five units in order of both higher
and lower speeds were extracted as shown in Fig. 5.

Fig. 7 Crack index in unit 13 and unit 73
Here, the values were evaluated by the cracking
level needed to be repaired. The period to reach thelevel A was 39 months for the unit 13 which was themost vulnerable to deterioration, on the other hand,that of the least vulnerable unit 73 was 585 months. It
was revealed that the period of deterioration was
remarkably different depending on units. The factor
to cause the difference in degradation over time wasinvestigated through the data of Kumamoto Airport.
Although the degradation speed of Kumamoto
Airport was able to be determined using the existing
maintenance data, it was difficult for a new or
repaired runway to derive its degradation speed dueto the lack of existing data. Therefore, a method topredict degradation speed was attempted to beestablished, where degradation speed was obtained by
calculating a fatigue coefficient considering

Fig. 5 Five units in order of both higher and lower
crack deterioration speeds
It was confirmed that the most crack-prone unit
was 13 and the least crack-prone units were 68 and 73
in the runway of Kumamoto Airport. Fig. 6 shows the
runway distribution of the crack degradation speed.
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GEOMATE- Bangkok, Nov. 14-16, 2016

pavement fatigue as a cracking factor of the asphalt
paving.

The lower a temperature becomes, the more
frequently cracks tends to occur, because a cracking
deformation coefficient becomes greater as apavement temperature is lower. The asphalt pavement
temperature was referred to the 2010 temperature in
Mashiki city where Kumamoto Airport [3] waslocated. Fig. 9 shows a 24-hour change oftemperature recorded in Mashiki city.

METHOD OF CALCULATING A
DEFORMATION COEFFICIENT OF
ASPHALT MIXTURE
First of all, a fatigue coefficient was calculated
based on the data of Kumamoto Airport. The
coefficient is defined as the value obtained by
dividing the fatigue degree of various kinds ofaircrafts by the fatigue degree determined under thecondition that the nation’s largest aircraft: B777-300
constantly took off and landed on the asphalt
pavement 100 times a month at a low speed and atemperature of 20 degrees Celsius. This coefficient isused for the airport pavement design for fatigue
cracking [2]. Formulas for calculating the degree of
fatigue are shown below.

D

Tv(i)
N f(i)(3)

1  1 
N f         E 

Fig. 9 24-hour temperature change from January to
December in Mashiki.

(4)The change of temperature between 7:40 and 21:00
was focused on because it was the time zone for the
take-off and landing in Kumamoto Airport. It was
cleared that takeoffs and landings around 8 and 21
o'clock in January had a great influence. The aircrafts
which mainly takeoff and land on the runway in
Kumamoto Airport are as follows: B767-300, B787,
B737-800, B737-500, B737-200, A320, DHC8-400,
Canadair Regional Jet 200, Embraer 170 and ATR42600. A deformation coefficient was calculated usinga pavement temperature at the take-off and landingtime of each aircraft. Fig. 10 shows the change inmodification coefficients by loading time when the
pavement temperature is constant.

where D: the degree of fatigue caused by aircraft,
Tv: the traffic volume of aircraft (i) (times), Nf: thenumber of repetitions of up to destruction (times), ε:
the tensile strain generated in asphalt mixture, E: thedeformation coefficient of asphalt mixture (MPa), α,β, γ: coefficients. A deformation coefficient changes
according to a temperature of pavement and loading
time of an aircraft on asphalt pavement. Since theloading time is the time period during which wheelsof an aircraft contact with the pavement, it tends todecrease as a speed of the aircraft increased. Fig. 8
shows the change in deformation coefficients due tothe pavement temperature when the speed of anaircraft is constant.

Fig. 10 Change in the modification coefficient byloading time when pavement temperature is constant.

Fig. 8 Change in deformation coefficient due topavement temperature when the speed of an aircraft
was constant.
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GEOMATE- Bangkok, Nov. 14-16, 2016

Since a deformation coefficient increases as aspeed of aircraft increases, cracking is likely to occur
more often. The velocity distribution of the B767-300
is shown in Fig. 11 as an example.

Fig. 12 Load distribution regarding
lift of aircraft B767-300
The speeds of take-off and landing were set at
333km / h and 270km / h, respectively. Here, theaircraft was supposed to enter the nearest taxiway
once the running speed has reduced to the speed equal
to a traveling speed on the taxiway at the time oflanding. In Kumamoto Airport, the unit 1 is usuallyused the most frequently for take-off and landing,however, the unit 100 located in opposition to the unit
1 is also used more frequently during the summer
under the influence of the change of wind direction.
Assuming these factors greatly affect the crack, an
expression of a fatigue coefficient was obtained
taking into consideration of these factors. An
expression that evaluates a fatigue coefficient isshown below.

Fig. 11 Velocity distribution of B767-300
As takeoffs and landings were supposed to be
performed from both ends of the runway, four
velocity distributions were made for a single aircraft.
The cracks of runway are significantly affected at thetime of take-off at the fastest speed of an aircraft.
Consequently, it was confirmed that a temperature of
the pavement and a velocity of an aircraft were
important factors to determine a deformation
coefficient.
METHOD OF CALCULATING A FATIGUE
COEFFICIENT F  x  E  y N n DID           k f D0
 F0   E0  N 0 n0 

To determine a fatigue coefficient, a tensile strainof asphalt pavements was used in the equation (4).
Since a tensile strain hardly changed depending on
the size of an aircraft, the variation of strains wasconverted into the load of the main landing gear andthe passing number of wheels. Since each aircraft has
different number of main landing gear and therefore
the passing number of wheels naturally becomes large
in the case of the aircraft with more wheels, thenumber of repetitions was utilized for the equation as
traffic volume. The weights of aircrafts werecompared by the load of the main landing gear which
supports 90% of the weight of an aircraft. Lift of anaircraft is enhanced according as the running speed
increases and accordingly the load of an aircraft
applied to the pavement may change [4]. Fig.12
shows the load distribution regarding the lift of theaircraft B767-300.

(5)where ID: the fatigue coefficient, D: the total of
fatigue rate of flying aircraft, D0: fatigue of B777-300 ,F: the main landing gear load of flying aircraft (kN),
F0: the main landing gear load of B777-300 duringrunning at 32km / h (kN), E: the deformation
coefficient of asphalt mixture (MPa), E0: the modulus
of deformation at the time of the pavement
temperature 20 degrees Celsius, N: the traffic volume
of a flying aircraft(times), N0: 2400 times, n: thenumber of wheel axis of B777-300, n0: the number ofwheel axis of a flying aircraft, kf: the proportion of
take-off and landing direction, x,y：1.9, 4.3. Fig. 13
shows the distribution of a fatigue coefficient inrunway location. Fig. 14 shows the correlation
between a fatigue coefficient and crazing degradation
speed.

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aircraft, a gliding distance, temperature of each
airport, wind direction and take-off and landing
behaviors.
CONCLUSIONSThis study examined the deterioration prediction of
cracking rate in Kumamoto Airport. As a result, thefollowing views were obtained.
1.

It is possible to determine cracking rate by fixing
non-linear increase of cracking rate to linear
form by power number and comparing theobtained slopes. The slope was named crazing
degradation speed.

2.

From the results, it was confirmed that crazing
degradation speed tended to be high in thevicinity of the unit 20 and low in the vicinity of
the unit 70 in Kumamoto Airport．

3.

It is possible to evaluate crazing degradation
speed using an expression of fatigue degreeincorporating load of aircraft, a gliding distance,
temperature of each airport, wind direction andthe take-off and landing behaviors.

4.

It was revealed not only that the fatigue
coefficient obtained through this study had a highcorrelation with crazing degradation speed, butalso that the cracking rate greatly affected a
deformation coefficient and the load in
Kumamoto airport.

Fig. 13 Comparison between a fatigue coefficient andcrazing degradation speed

REFERENCES[1] Ministry of Land, Statistics information. 2014.
December. 12 <http://www.mlit.go.jp/statistics/
details/cab_list.html>

Fig. 14 Correlation between a fatigue coefficient andcrazing degradation speed.
The difference between a fatigue coefficient andcrazing degradation speed became large at points near
the unit 60 as seen in Fig. 13. These diverged plots
are assumed to be caused by a mistake of inspection
because crack rate reduced with time elapse eventhough any repair work has not been performed. As
shown in Fig. 14, although the measured values in thearea surrounded by a red circle has low reliability, thecorrelation between a fatigue coefficient and crazing
degradation speed was confirmed except for a little
difference. Moreover, it was found out that it waspossible to evaluate crazing degradation speed by theexpression of fatigue degree incorporating load of[2] Ministry of Land. Japan Civil Aviation Bureau.
2008. Airport pavement design point and design
example,2008.ppI10-I13.
[3] Ministry of Land, The Meteorological Agency.
2014.November.30. <http://www.data .jma.
go.jp/obd/stats/etrn>[4] Formula to calculate the lift of the wing.2014.
October, 30 <http://www.cfijapan.com/study/
html/to099/html-to035/017b-Lift_CL.htm>

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ARTIFICIAL NEURAL NETWORK PERMEABILITY MODELING OF
SOIL BLENDED WITH FLY ASH
1,2

Joenel G. Galupino1 and Jonathan R. Dungca2
Faculty, De La Salle University, Manila, Philippines

ABSTRACTThe determination of the permeability properties of soil is important in designing civil engineering projects
where the flow of water through soil is a concern. ASTM D2434 Standard Test Method for Permeability ofGranular Soils (Constant Head & Falling Head) is being followed to determine the vertical permeability, while
for horizontal permeability, there are none. In this study, tests such as Atterberg limit, relative density tests, andparticle size analyses are done to determine the index properties of soil blended with fly ash. Subsequently,
microscopic characterizations tests, elemental composition tests and permeability tests are done to determine thechemical and physical properties of the soil mixes. A new permeability set-up was used in determining thehorizontal permeability soil mixes. Data were extracted during the experiment and a relationship between theproperties of soil and the permeability was established. An artificial neural network model was used to predictthe coefficient of permeability when the percentage of fly ash is available.
Keywords: Permeability, artificial neural network, modeling, fly ash, waste utilization

INTRODUCTIONMETHODOLOGY

It is very important for engineers to understandthe soil underneath because it will affect the way the
structures are designed. Without its knowledge, lives
would be at stake and it can reveal a lot of
information by just analyzing the soil profile of acertain area.
Geotechnical properties of the soil (grain size
distribution, Atterberg limits, specific gravity,
maximum and minimum index densities, soil
classification, permeability, shear strength andcompressibility) should be considered by engineerswhen designing foundation, retaining walls, etc. butthese properties change over time and are influenced
by physical content, climate and weather.
Permeability generally relates to the propensityof a soil to allow fluid to move through its void
spaces [1] and is vital to every project where theflow of water through soil is a concern.
Coal-fired power plants discharge large amountsof fly ash as waste but only half of them are used
and the remaining half is trashed to land and sea, its
disposal became an environmental concern, there arealso local studies in the Philippines that shows theeffect of wastes to the environment [2,3,4]. The
utilization of fly ash may be a viable alternative forporous backfill material because fly ashes generally
consist of silt-sized particles and consequently
possess high permeability [5] but tests must be done
to determine the permeability [6] of soil-fly ash
mixture since there was a lack of information on thehorizontal permeability of the said mixes.

Fly ashes generally consist of silt-sized particles
and consequently possess high permeability as
mentioned by Prashanth (2001) [5].
Shown on Table 1 are the soil mixtures that werechecked on the effect of fly ash on soil.
Table 1. Soil Mixtures
Fly AshSoil Mixture
(%)
100FA
100
75FA25S
75
50FA50S
50
25FA75S
25
100S
0

Soil
(%)
0
25
50
75
100

The density of the soil mixtures were determinedusing ASTM D854 [7], which is the standard for the
specific gravity tests. In determining the Liquid
Limit, Plastic Limit and the Plasticity Index of thesoil mixtures, Atterberg limit tests based on ASTM
D4318 was used [8]. Furthermore, the maximum(e max ) and minimum (e min ) index densities for soil
mixtures was discerned using the ASTM D4253 [9]and ASTM D4254 [10], respectively. Then theparticle size analyses using ASTM D422 wasutilized to determine the percentage of different
grain sizes in a soil [11].
Scanning electron microscopy (SEM) withenergy dispersive X-ray spectroscopy (SEM/EDX)
is the best known of the surface analyticaltechniques for different materials. The scanningelectron microscopy (SEM) was used to evaluate themicroscopic characterization of each soil mixture
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GEOMATE- Bangkok, Nov. 14-16, 2016

which produces high resolution images of surface
topography, while Energy Dispersive X-ray
Spectroscopy (EDX is used to determine thechemical composition of the soil and gives
information on the elements present in the soil
mixture.
The permeability of the different soil mixes were
determined by the constant head test method andfalling head test method. The direction of the flow
weas also considered, thus, vertical and horizontal
orientations of permeameter were used. A proposed
set-up by Smith (2010) [12] for permeameter wasused and modified to determine the horizontal
permeability of the soil mixtures, shown on Fig. 1.
The equation utilized for the permeability set-up is
Eq. 1.

RESULTS AND DISCUSSIONS
Index Properties
Using ASTM D854 the specific gravity of eachsoil blend was determined. The specific gravity ofthe soil mixtures was reduced by the addition of fly
ash [14] since the usual of the specific gravity of fly
ash is much lower compared with the soil. Shown on
Table 2, we can complement the study of Prabakar(2004) that the addition of fly ash (due to the lightweight property of fly ash) reduces the specificgravity of a soil mixture.
Table 2. Summary of Specific Gravity
Soil Mixture
Gs
100FA
2.02
75FA25S
2.11
50FA50S
2.31
25FA75S
2.49
100S
2.58

Manometer

Air-Water
CylinderPermeameter
Outflow
Collector

Pump

Table 3. Summary of Atterberg Limits
Soil
LL
PL
Mixture100FA
66
65
75FA25S
64
57
50FA50S
61
49
25FA75S
59
45
100S
52
32

Fig. 1. Horizontal Permeability Set-up
𝒌𝒌 =

𝑸𝑸𝑸𝑸
𝑨𝑨𝑨𝑨𝑨𝑨Fly ash is considered as silt material, it isexpected to have a plasticity index less than 1 basedon stablished literatures [14], thus, by adding fly ash
in the mixture has reduced the plasticity of a soil
mixture. Results are shown in Table 3.

(Eq. 1)

where:
k = coefficient of permeability, cm/s;
Q = quantity (volume) of water discharged
during test, cm3;l= length between manometer outlets, cm;
A = cross-sectional area of specimen, cm2;h = head (difference in manometer levels)
during test, cm;t= time required for quantity Q to bedischarged during test, s.

PI
1
7
12
14
20

It can be noticed from Table 4, the Maximum
Void Ratio (e max ) ranges from 1.78 to 1.99 because
the fine contents of the fly ash contributed to thepercentage of voids. 100S has the lowest value while
100FA has the highest, also from Table 4, 100S has
the lowest fines content, while 100FA garners thehighest. Their fines content and microfabric may
have contributed to the minimum and maximum
void ratio. These minimum and maximum void
ratios together with the target relative density of
90% were used to determine the void ratio to be
utilized for the permeability specimens.
To determine the maximum and minimum void
ratios of the different soil-fly ash mixes ASTM
D4253 and ASTM D4254 were used.
100FA has the greatest percentage of finescompared with other mixtures. Fly ash and soil areconsidered fines but the classification differ, fly ash
is silt and soil is plastic. It can also be noticed thatwhen fly ash is mixed with other soils increases the

Many numerical modeling techniques have been
introduced in the current technological era, and oneof them is Artificial Neural Network (ANN).
Artificial Neural Network can handle non-linear
relationships between variables and incomplete data
sets [13]. The proposed Artificial Neural Network
Model will be validated by involving a 45-degreeline as a guideline that provides insight into themeasured variables and as a critical part of theanalysis.

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fines content. The summary of results from theparticle size analyses are shown on Table 5.
Table 3. Summary of e min and e max
Soil Mixture
e min
100FA
0.27
75FA25S
0.37
50FA50S
0.47
25FA75S
0.72
100S
0.84

images of surface topography of the different soil
mixtures were produced using these tests. Pure soiland Fly Ash were initially tested to check theirmicroscopic characteristics, mixed soils were also
tested thereafter.
A combination of extremely strandy grains, large
angular grains and abundant silt grains formed themicro fabric of 100S. It is well-gradedmicroscopically and the silt grains have a roughsurface. A smaller inter-particle voids were createdby the smaller particles tend to fill the voids createdby the larger particles shown in Figure 1. Looking
closer to magnification of 1000x and 5000x, strandlike particles are present, his indicates that theseelongated particles also fill the voids, giving small
passageways for water to permeate.
While for 100F, it is a combination of larger silt
grains and smaller silt grains to form the micro
fabric normally 0.002-0.05 mm in size. Comparedwith silty sand (soil), particles have almost similarsize, forming larger inter-particle void that allows
water to pass through. On the 1000x and 5000x
magnification, the surface of the particle is not
smooth, this create passageway/voids for water topass through.

e max
1.99
1.98
1.94
1.93
1.78

Table 5. Summary of Particle Size Analysis Results%
Soil
Passing
D 10
D 30
D 60
Blend
#200
100F
61.83
0.029
0.03
0.04
75FA25S
50.78
0.019
0.032
0.06
50FA50S
29.79
0.032
0.0375
0.12
25FA75S
25.79
0.015
0.042
0.15
100S
21.84
0.01
0.4
1.2
SEM/EDX Results
In the Energy Dispersive X-ray Spectroscopy(EDX), chemical composition of soil is determinedto give information on the element present in the soil,
shown in Table 6. Oxygen (O) is very abundant,
followed by Silicon (for Silty Sand) and Calcium(for Fly Ash). Silicon and Calcium are predominant
in the soil elemental composition. Due to thepresence of Oxygen and other dominant elements:
Silica (from Silicon), Lime (from Calcium) andAlumina (from Aluminum) are the dominant
minerals
in
the
soil
sample.

Fig. 1. Microfabric of 100S (5000x, 1000x and 500x
Magnification)

Table 6. Summary of Elemental Composition
Composition
Composition(%) for Silty
Element(%) for Fly Ash
Sand
C, Carbon
17.39
5.41
O, Oxygen
46.65
40.64
Al, Aluminum
11.52
5.26
Si, Silicon
15.63
9.1
K, Potassium
1.05
0.78
Ca, Calcium
0.24
21.82
Fe, Iron
5.72
16.34
Cu, Copper
1.8
0.26
S, Sulfur
0
0.39

Fig. 2. Microfabric of 100FA (5000x, 1000x and
500x Magnification)

Fig. 3. Microfabric of 50FA50S (5000x, 1000x and
500x Magnification)

Most of the soil properties and characteristics
like strength, compressibility and permeability are
ascribed by its microfabric or microstructure. To
evaluate the microfabric of soil, fly ash, the scanningelectron microscopy (SEM) was used. Scanningelectron microscopy (SEM) with energy dispersive
X-ray spectroscopy (SEM/EDX) is the best knownof the surface analytical techniques. High resolution

A combination of extremely strandy grains,large angular grains and abundant larger silt grains
and smaller silt grains formed the micro fabric of
50FA50S. Looking closer to magnification of 1000x
and 5000x, strand-like particles are present but noy
prevalent compared with the pure soil, the soil
508

GEOMATE- Bangkok, Nov. 14-16, 2016

particles may contribute to the reduction of
permeability but the silt grains of fly ash willcounteract to allow water to drain faster.

head permeability test. To determine the effect of fly
ash in the horizontal permeability when added to soil,a box and whisker plot was drawn and shown on
Figure 5. The data garnered in this study agree withthe study of Das (2008) [16], that the horizontal
permeability values are larger than the verticalpermeability values, this is due to the pressure head
induced during the permeability test. The specimen
is laid in a horizontal position, which experiences nopressure drop within its body, unlike the verticalspecimen, which experiences pressure drop,
resulting to a slower flow of water.

Permeability
The study of Smith (2010) [15] and was
modified to determine the horizontal permeability ofthe soil mixtures. Shown in Table 7, are the range ofpermeability values gathered for the vertical oriented
constant head permeability test, to determine theeffect of fly ash when added to soil, a box andwhisker plot is delineated, shown on Figure 4.
A proposed approach in determining the verticalpermeability of the various soil mixtures was
utilized, it was referred on. The results of theexperiment agrees with the study of Prashanth(2001) that fly ashes generally consist of silt-sizedparticles and consequently possess high permeability
since it is prevalent that the permeability is increased
when the amount of fly ash is increased. Thus, theamount of fly ash increase the permeability of thesoil mixes.

Table 8. Range of permeability values for horizontal
oriented permeability test
Soil
Minimum k,
Maximum k,
Mixture
cm/s
cm/s
100FA
6.02E-05
7.28E-05
75FA25S

4.25E-05

5.02E-05

50FA50S

3.40E-05

4.04E-05

25FA75S

3.04E-05

3.70E-05

100S

2.21E-05

2.70E-05

Table 7. Range of permeability values for verticaloriented permeability test
Soil
Minimum k,
Maximum k,
Mixture
cm/s
cm/s
100FA
4.53E-05
5.52E-05
75FA25S

3.40E-05

3.80E-05

50FA50S

2.55E-05

3.16E-05

25FA75S

2.05E-05

2.51E-05

100S

1.47E-05

2.09E-05

100FA 75FA25S 50FA50S 25FA75S 100S

Soil Mix

Fig. 5. Effect of fly ash on the horizontal
permeability when added to soil
Fly ash is the recommended addition to the soil
mixtures since waste materials are aimed to be
utilized and the addition of fly ash, which has a
combination of larger silt grains and smaller silt
grains to form the micro fabric prevalent to themicroscopic characterization test for 100F, to soils
changes the inter-particle void ratio [14]Subsequently, silt particles have almost similarsize, forming larger inter-particle void, contributing
to a much larger inter-particle voids, thus, thepermeability of pure fly-ash ranges: (1) verticaloriented 4.51x10-05 cm/s to 5.35x10-05 cm/s and (2)horizontal oriented 1.93x10-05 cm/s to 7.29x10-05
cm/s.

100FA 75FA25S 50FA50S 25FA75S 100S

Soil Mix
Fig. 4. Effect of fly ash on the vertical permeability
when added to soil

In determining how long the contaminated water
will penetrate the ground water, the horizontal
permeability of soil mixtures shall be considered.
Shown in Table 8, are the range of permeability
values gathered for the horizontal oriented constant509

GEOMATE- Bangkok, Nov. 14-16, 2016

75FA25S, 50FA50S, 25FA75S, 96S4FA are themixtures that include fly ash and soil, theirmicrofabric is a combination of extremely strandy
grains, large angular grains and abundant larger
rough-surfaced silt grains and smaller roughsurfaced silt grains. Evidently shown on Figure 4
and Figure 5, as the amount of fly ash is increased,
the drainage also increased. Due to the contributionof fly ash to the inter-particle voids of the soil
mixtures, the permeability of mixture of soil and flyash ranges: (1) vertical oriented 1.93x10-05 cm/s to
3.80x10-05 cm/s and (2) horizontal oriented 2.52x1005
cm/s to 5.02x10-05 cm/s.
To validate the results of the vertical oriented
and the horizontal oriented permeability tests, theirratio must be within the given range of Das (2008).
The collected usual ratio of horizontal and verticalpermeability of soils by Das (2008) is with the rangeof 1.2-3.3, thus, the data gathered are between 1.31.5, thus, ratios are within Das’ desired range.

By determining the number of neurons in the
input and output layers, number of hidden layers andthe number of neurons in each hidden layer, the best
Artificial Neural Network model for the soil
mixtures can be garnered. Thus, after numerous trial
the ANN structure 5-9-2 (5 input, 9-nodes and 2
output) was determined to be the best model toestimate the permeability of the soil mixtures. The
model was able to give an acceptable values of R:
0.98414, 0.97570, and 0.96843 for validation,
training and testing, respectively. Figure 6,
represents the regression line for ANN 5-9-2. Also,
the 45deg line shows a validation and an agreement
between the experimental (actual) and the predictedparameters. Subsequently, Figure 7 shows theperformance curve of the ANN model, the meansquared error is 0.0061175 and occurred at epoch 25.
Best Validation Performance is0.0061175 at epoch 25

Best Validation Performance is 0.0061175 at epoch 25
10

0

Train

Artificial Neural Network

Validation
Test
Best

In the Artificial Neural Network Model, five (5)variables were considered:
1. The percentage of added fly ash in the soil
2. Specific gravity
3. Liquid Limit
4. Maximum Void Ratio
5. Minimum Void Ratio
The data garnered were divided into three (3)groups: 70% for training the neural network, 15%
for validation and 15% for testing. In the hidden
layer, tan-sigmoid transformation function was
utilized, while in the output layer a linear
transformation function was used. The feed-forward
backpropagation technique was used to generate thebest model for estimating the permeability of the soil
mixtures. Also, fastest backpropagation algorithm
and highly recommended Levenberg-Marquardt
network training function was also employed. The
authors utilized Matlab for the ANN Algorithm.

Mean Squared Error (mse)

10

10

10

-1

-2

-3

0

5

10

15

20

25

30

31 Epochs

Fig. 7. Best validation performance of the model
There are many studies that provided a
numerical approach in determining the permeability
when the amount of fly ash is available [17]. The
permeability coefficient was further compared witha regression model. The said regression model gavean R2 value of 0.9335 which is lower compared to
calculated R2 value of the ANN method forvalidation 0.9685, 0.9520, and 0.9379 for validation,
training and testing, respectively.

Regression Line
-9.5

Output=~0.97*Target-0.33

Data

CONCLUSION AND RECOMMENDATIONS

Fit
Y=T

The addition of fly ash to soils changes the interparticle void ratio [5], it increases the permeability,
thus, the microscopic characteristics of the soil
mixtures may contribute to the increase inpermeability. Based on the tests, fly ash is acombination of larger silt grains and smaller silt
grains to form the micro fabric. Silt particles havealmost similar size, forming larger inter-particle void,
contributing to a much larger inter-particle voids.
ANN Structure 5-9-2 is the best architecturesince it has the highest correlation coefficient, R2,
and the said model has a good predicting ability

-10

-10.5

-11

-11

-10

-10.5

-9.5

TargetFig 6. Regression Line for ANN Structure 5-9-2
510

GEOMATE- Bangkok, Nov. 14-16, 2016[7] American Society for Testing and Materials.
(n.d.). Standard Test Methods for Specific
Gravity of Soil Solids by Water Pycnometer.
ASTM D854. ASTM D854.
[8] American Society for Testing and Materials.
(n.d.). Standard Test Methods for Liquid Limit,
Plastic Limit, and Plasticity Index of Soils.
ASTM D4318.
[9] American Society for Testing and Materials.
(n.d.). Standard Test Methods for Maximum
Index Density and Unit Weight of Soils Using aVibratory Table. ASTM D4253.
[10] American Society for Testing and Materials.
(n.d.). Standard Test Methods for Minimum
Index Density and Unit Weight of Soils andCalculation of Relative Density. ASTM D4254[11] American Society for Testing and Materials.
(n.d.). Standard Test Method for Particle-SizeAnalysis of Soils. ASTM D422.
[12] Smith, C. B. (2010). Horizontal Permeameter.
Magna Cum Laude Honors Research Project,
University of Florida, Department of Civil andCoastal Engineering, Gainesville, Florida.
[13] Oreta, A., & Ongpeng, J. (2011), “Modeling theconfined compressive strength of hybrid
circular concrete columns using neuralnetworks”, Computers and Concrete, 8(5), 597616.
[14] Prabakar, J., Dendorkar, N., & Morchhale, R.
(2004). Influence of fly ash on strength behaviorof typical soils. Construction and Building
Materials, 18, 263-267.
[15] Smith, C. B. (2010). Horizontal Permeameter.
Magna Cum Laude Honors Research Project,
University of Florida, Department of Civil andCoastal Engineering, Gainesville, Florida.
[16] Das, B. M. (2008). Advanced soil mechanics.
New York: Taylor & Francis.
[17] Galupino, J., Dungca, J. (2015). Permeability
characteristics of soil-fly ash mix. ARPN
Journal of Engineering and Applied Sciences,15, 6440-6447.

compared with previous studies. To improve theresults, extensive manipulation of parameters should
also be considered. Sensitivity analyses may be
conducted to further validate the model and theresults.
Furthermore, it is recommended for the purposeof ground improvement engineering, testing theshear strength and the compressibility of the soil
mixtures should be determined and relate them tothe permeability.
REFERENCES[1] Liu, C., & Evett, J. B. (2000). Soil Properties:
Testing, Measurement and Evaluation. (E.
Francis, Ed.) Charlotte, North Carolina, USA:
Prentice Hall.
[2] Caplano, J. (2001, December 30). Closed
landfills contaminating groundwater, WorldBank warns. (Philippine Star) Retrieved
November 2013, from Philippine Star:
http://www.philstar.com/nation/145333/closedlandfills-contaminating-groundwater-worldbank-warnsAnnnn B, “Unpublished work but
accepted”, accepted, Year.
[3] World Bank. (2012, October 24). Philippines:
Toward Greener Waste Management. Retrieved
November
2013,
from
World
Bank:
http://www.worldbank.org/en/news/pressrelease/2012/10/24/philippines-toward-greenerwaste-management[4] Adajar, M. A., & Zarco, M. (2014, September).
An empirical model for predicting hydraulicconductivity of mine tailings. International
journalof
GEOMATE:
geotechnique,construction materials and environment. 7,1054-1061.
[5] Prashanth, J., Sivapullaiah, P., & Sridharan, A.
(2001). Pozzolanic fly ash as a hydraulic barrierin landfills. Engineering Geology, 60, 245-252.
[6] Baxter, D. Y. (2000, April 14). Mechanicalbehavior of soil-bentonite cut-off wall. Virginia
Polytechnic Institute and Stat University,
Department of Civil Engineering. Blacksburg,
Va.

511

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ITERATIVE ALGORITHM TO CONSTRUCT AN EXACT FINITEELEMENT MODEL FOR AXIALLY LOADED PILE IN ELASTOPLASTIC SOIL
Chinapat Buachart and Chayanon Hansapinyo
Department of Civil Engineering, Faculty of Engineering, Chiang Mai University

ABSTRACT
An iterative algorithm to construct a displacement based finite element method for analyzing axially loaded
pile embedded in finite depth of elasto-plastic soil is presented. The investigation herein is conducted on thecondition of shape function by which exact solutions may be reproduced at the nodal points regarding to a few
number of elements. The examined shape functions which satisfy the homogeneous governing equations inelastic and plastic soil through bisection iterative algorithm are introduced to obtain the so-called exact element
stiffness matrix via total potential energy principle. Numerical examples of elastostatic pile embedded in elastoplastic Winkler foundation illustrate the accuracy of proposed element compare with conventional finite element
shape functions. Axial force and displacement solutions show very good agreement with data from the available
literature i.e. the exact nodal displacement solution is obtained correspond to any point load level even with asingle element mesh employed.
Keywords: Axially loaded pile, Displacement method, Finite element, Iterative algorithm, Soil-pile interaction

resistance will keep constant as long as thedisplacement increases. The soil bearing capacity at
pile’s end is presented by coefficient k b . The pile isassumed to behave as an elastic column with
Young’s modulus E. The Poisson’s ratio of the pilematerial is neglected. Figure 1(a) shows the elastic
and plastic zone occurred in soil due to the axial
displacement w 0 at top pile head is greater thanyielding displacement w * . Suppose that yielding
displacement occurred at depth z 0 from pile head.
Hence, the length of an elastic portion in Fig. 1 isdenoted by ℓ, the magnitude of axial displacement inelastic portion is less than or equal to yielding
displacement.

INTRODUCTION
In this study, the displacement of axially loaded
pile embedded in elasto-plastic soil is solved via
proposed finite element procedure. The nodal exact
shape function concept suggested in [1]–[4] are used
to construct the tangent stiffness matrix andequivalent nodal force incorporate with NewtonRapshon algorithm to solve nonlinear algebraic
equations. An internal point which is separate elastic
and plastic portion soil around embedded pile thendetermined via simple iterative bisection procedure.
Example of elasto-static pile embedded in elastoplastic soil subjected to quasi-static point load on top
soil level is analyzed. The results from proposed
element are compared with analytical solution from
[5] and conventional simplex bar element to verify
the accuracy of proposed pile element.

R

R

R

R

R

R

Consider one-dimensional element in Fig. 2.
Assuming that soil surrounding pile element is inelastic condition for whole length. The total
potential energy of this soil-pile element subjected tothe axial forces P 1 and P 2 is defined as the sum ofinternal potential energy (strain energy) and theexternal potential energy due to external load as
follow [4]:Problem Definition

The analysis considers a single circular pile [4],with diameter d (cross section A = 0.25πd2 andperimeter U = πd), embedded into soil deposit (Fig.
2
LL1). The pile has a total length L and is subjected to
1 dw 
withan axial force P 0 at the pile head which is flush=
P
EA
dz
+
k
U
w2 dz  ∫ 
s∫2  0  dz 
the ground surface. The soil medium is assumed to0
be elastio-plastic, isotropic and homogeneous, with− Pw
− P2 w2
1 1
elastic properties described by equivalent springcoefficient k s . Once the soil displacement goesbeyond the yielding displacement w * , the shear
R

R

R

R

R

Governing Differential Equations

MATHEMATICAL FORMULATION

R

R

R

512

(1)

GEOMATE- Bangkok, Nov. 14-16, 2016

where α2 = k s U /EA is the characteristic parameter ofpile surrounding by elastic soil region.

where w(z) is the vertical pile displacement at depthz. The first and second terms in Eq. (1) represent thestrain energy in pile and surrounding soil,
respectively.

P0

W0plastic zone

Z0●

elastic zoneℓFig. 2

Axially loaded pile element [4]A set of natural boundary conditions are as follows

kb

dw
dw
(4)
P1 =
− EA
and P2 =EA
dz z 0=dz z L=(a)stressPlastic soil conditions
Suppose that ended load at top soil level reachsome limit, the soil portion at level 0 < z < z 0 is inplastic condition. Consider the plastic soil portion,
upper portion in Fig. 1, displacement w in second
term of equilibrium equation (3) have to be replacedby yielding displacement w * as follow

fs

1

ks

2

fs
−=
0 for 0 < z < z0
EA
dz

d w

W*

2

displacement

In which f s = k s Uw * is the maximum shearing force
occurred in soil surrounding pile element. Shearing
force in soil surrounding pile element is linearly
dependent on pile displacement until they reachsome limit of yielding displacement magnitude w * .
The nodal exact finite element solutions of Eqs.
(3) and (5) for a given value of plastic soil depth z 0
can be found in [4]. In present work, the proposed
finite element procedure to solve Eqs. (3) and (5) fora given load P 0 (Fig. 1) will be explained in next
section. Then, the results from proposed nodal exact
element and conventional linear element will becompared with available analytical solution in [5].

(b)Fig. 1

Axially loaded pile and soil model [4]The first variation of Eq. (1) leads to d d w  dw dz + k U ( d w )wdz
s
∫  dz ∫ dz 00
L

dP EA =L(2)

− P1d w1 − P2d w2Applying the appropriate Gauss-Green theorem to
Eq. (2) and setting δП = 0, gives the differential
equation for equilibrium
2

d w
dz

2
−α
=
w 0 for 0 < z < L
2(5)

Newton-Raphson Iteration
Due to elasto-plastic behavior of shear force insoil surrounding pile element, finite element
discretization of Eqs. (3) and (5) leads to the steadysteady-state set of non-linear algebraic equations
given by the residual equation as follows [6]–[9]:

(3)

513

GEOMATE- Bangkok, Nov. 14-16, 2016

r = f − F (w
(k )

(k )

)

in Fig. 1(a). The displacement of soil surrounding
pile element still be in elastic condition, .i.e. w(z) ≤w * . Hence, trial solution of pile in elastic soil
condition was in hyperbolic functions as describedin [4]. The internal force of proposed element can be
expressed as follow:(6)

In which internal force F is a non-linear function(bilinear function in this case) of nodal displacement
w at iteration number k. The external nodal load f ispresented in term of specified vector. A solution ofresidual r = 0 is obtained by solving an algebraic
equation:

K T ∆w ( k ) =
r( )
(k )

k

 coth β
F = α EA 
 −cschβ

Conventional linear element
To compare the accuracy and efficiency of proposed
element, the internal force of conventional linear
finite element has to be recalled. An expression ofinternal force vector for conventional linear finite
element in elasto-plastic soil condition is shown in
Eq. (11).

(8)

An iteration is repeated until the magnitude ofincremental nodal solution Δw and residual r close
to zeros.

F=

Element Internal Force
Nodal internal force F in Eq. (6) can be derived
from first variation of strain energy term in Eq. (2).
In conventional linear finite element, the trial
solution w(z) and variation δw(z) are approximatedvia linear shape functions N a (z), a = 1 or 2, which
results in non-exact nodal solution w. To produce
the nodal exact solution of unknown w, the trial
solution w(z) and variation δw(z) are approximated
via the set of shape function N a (z) which satisfy
homogeneous solutions of governing equations (3)and (5) as discussed in previous work [4]. Hence,
nodal internal force of conventional and proposed
two nodes element can be expressed as follow.

+
+

2 3
 2 ( L + 2 z0 )   w1 
α 2 EA  (11)
 2
3
3 
6 L2  ( L + 2 z0 ) 2 ( L − z0 )   w2 
f s  z0 ( L +  ) 



2L 

z02




Tangential Stiffness Matrix
Tangential stiffness matrix in Eq. (7) will be
constructed from matrices of internal force in Eqs.
(9) and (10). First, nodal internal forces from Eq. (9)and (10) are assembled to form three members
vector, following degree of freedom w 1 , w 2 and w * .
Then degree of freedom corresponding to yielding
displacement w * will be condensed out via staticcondensation technique [7]. Hence, tangential
stiffness matrix is reduced stiffness represents a setof two equations with two unknowns corresponding
to ended displacements, w 1 and w 2 .

Consider the upper part of soil (plastic zone) in
Fig. 1(a). The displacement of soil surrounding pileelement is in plastic condition, .i.e. w(z) ≥ w * . Theinternal force of proposed element can be expressed
as follow:

EA  1 −1  w1  f s z0 1 +
z0  −1 1   w*  2 1

−1  w1  
L  −1 1   w2 

EA  1

Then the procedure to construct tangent stiffness
matrix used to determine incremental displacement
following Eq. (7) will be explained.

Plastic soil conditions

F

(10)

In which non-dimensional parameter β = α  .

(7)where K T (k) = ∂F/∂w(k) is so-called tangential
stiffness matrix, used to obtain the incremental nodal
solution Δw. Then update the nodal solution:+1)w ( k=w ( k ) + ∆w ( k )−cschβ   w*  coth β   w2 

(9)Estimation of plastic zone
Note that the internal force Eq. (9) is derived fromlinear shape function as shown in appendix

Construction of the internal force term in Eqs.
(9)–(11) require the estimated length of plastic zone
(z 0 ) at present iteration. Let assume that the yieldingof soil surrounding pile element occurs in upperzone of pile element in Fig. 2, i.e. w 1 > w 2 and w 2 <Elastic soil conditions
Now consider the lower part of soil (elastic zone)
514

GEOMATE- Bangkok, Nov. 14-16, 2016w * < w 1 . Imply the continuity of internal axial force
via finite element equilibrium equation at z = z 0(ghost node). Hence, relation between yielding
displacement at given position z 0 and specified valueof nodal solutions (w 1 and w 2 ) for proposed nodal
exact element is:w* =w1 sinh β + w2α z0 (α z 0 ) 2 1 + sinh β + α z0 cosh β
2 proposed and conventional elements are shown in
Table 1. Finite element solutions of proposed
element are identical to the exact analytical solutions
in [5]. Additionally, the conventional linear element
behaves stiffer than proposed element. Pile headdisplacements of conventional linear element areless than results from proposed element andanalytical solution [5].

(12)
Table 1 Calculated plastic depth and settlement ofpile head at any values of load P 0 .
Present (Exact)
Linear FEM
P 0 (kN)
z0
w0
z0
w0
(m)
(mm)
(m)
(mm)
2.21
2086
0
2.60
3.24
2951
9
3.91
9.64
4.96
3796
18
5.64
23.60
7.29
4593
27
7.71
33.49
9.82
5291
36
9.95
40.37
11.98
5807
45
11.98
45

Equation (12) is numerically solved via bisection
method to obtain the length of plastic zone, namely
z 0 . Then the tangential stiffness matrix K T can beobtained from procedure described in previous
section. Note that if the continuity of force fromnatural boundary conditions, Eq. (7) were used toderive the relation between nodal displacements andlength of plastic zone instead of finite element
equilibrium equation. Then the result is similar with
Eq. (12), except that the square term (αz 0 )2 in
denominator of Eq. (12) will be disappeared.
Besides, in the case of conventional linear finite
element, the value of plastic length z 0 can be
estimated directly from linear interpolation, i.e.

 w −w 
z0 =  1 *  L w1 − w2 

For comparison, the load and settlement in Table
are also plotted in Fig. 1. The load-settlement
relations curves show nonlinear behavior when theload at pile head is gradually increased.

(13)

For a given value of point load at pile head, i.e.
force P 0 in Fig. 1(a), iteration processes in Eqs. (7)and (8) are repeated until equilibrium conditions
satisfied.
NUMERICAL EXAMPLESIn this section, numerical example is presented toillustrate the effectiveness of nodal exact finite
element proposed in previous section. Results fromproposed element then compare with conventional
linear finite element. Accuracy of proposed element
is verified using analytical solutions available in [5].
A bored pile was installed in the medium silt
clay and the end bearing layer is sandstone. The pilelength is 45 m, and the diameter d = 1 m. The elastic
modulus of pile shaft E = 2.2 × 107 kPa. From soil
tests, the values of equivalent soil elastic coefficientk s = 12000 kPa/m, the yielding displacement of soilw * = 2.6 mm, and end bearing stiffness k b = 684000
kPa/m [4]. The value k b A is added into the last
diagonal member of stiffness matrix.
The numerical test was performed using one
proposed nodal exact element and one conventional
linear element. The magnitude of point load at pilehead P 0 is gradually increased. The results from

Fig. 1

Load-settlement curves of proposed versus
conventional finite elements.

CONCLUSION
The finite element model for pile embedded inelasto-plastic soil subjected to axial load is proposed.
The nonlinear algebraic equation has been solved via
Newton-Raphson iteration. Tangential stiffness isconstructed through static condensation techniqueand bisection search method. Numerical example forstatic load pile embedded in elasto-plastic soil wastested by gradually increase load on pile head until
shear force in surrounding soil reach ultimate soil
capacity. Results obtain from proposed element werecompared with analytical solution and conventional
linear finite element solution. Numerical test
indicates that an exact finite element solution can be
515

GEOMATE- Bangkok, Nov. 14-16, 2016

obtained even with minimum number of elementused. Linear conventional finite element behaved
stiffer than the proposed nodal exact finite element.

tanh ( γ ) =REFERENCES

The ratio of displacement at the pile head w 0
with respect to yielding displacement is therefore[1] Tong P, “Exact solutions of certain problems byfinite element method”, AIAA J., Vol. 7, 1969,
pp. 178-180.
[2] Kanok-Nukulchai W, Dayawansa PH, andKarasudhi P, “An exact finite element model fordeep beams”, Int. J. of Struc, Vol. 1, Jan. 1981,
pp. 1-7.
[3] Ma H, “Exact solutions of axial vibration
problems of elastic bars”, Int. J. Numer. Meth.
Eng., Vol. 75, 2008, pp. 241-252.

kb
aE

(15)

(a z 0 )w0=1 + aaz0 tanh (  + γ ) +
2
w*

2

(16)

In addition, ratios of pile tip displacement w t withrespect to yielding displacement:cosh ( γ )wt=w* cosh (α  + γ )(17)[4] Buachart C, Hansapinyo C, and Sommanawat
W, “An exact displacement based finite element
On deriving Eqs. (14)–(17), soil under pile tip ismodel for axially loaded pile”, Int. J. ofassumed to be in an elastic condition. Detail
GEOMATE, Sept., Vol. 11(25), 2016, pp. 2474derivation of Eqs. (14)–(17) are described in
2479.
reference [5].
[5] Chou YC and Hsiung YM, “A normalized
equation of axially loaded piles in elasto-plastic
A.2 Shape Function of Nodal Exact Element [4]soil”, J. Geo Eng., Vol. 4(1), 2009, pp. 1-7.
[6] Bathe KJ, Finite Element Procedures. PrenticeIn reference [4], shape functions of proposed pile
Hall, 1996, ch. 6.
element were derived from homogeneous solutions[7] Cook RD, Malkus DS, Plesha ME, and Witt RJ,
of Eqs. (3) and (5). There are expressed in theConcept and Applications of Finite Element
following forms:
Analysis. Wiley & Sons, 2002, ch. 6 and 17.
[8] Simo JC and Hughes TJR, Computational
Plastic soil conditions (0 ≤ z ≤ z 0 )Inelasticity. New York: Springer, 1998, ch. 1.
[9] Zienkiewicz OC and Taylor RL, The Finite
z0 − zz
Element Method. Vol. I. Oxford: Butterworth=
N1 z=, N2 z(18)
Heinemann, 2000, ch. 20.
z0z0[10] Chapra SC, Applied Numerical Methods with
Matlab for Engineers and Scientists. McGrawElastic soil conditions (z 0 ≤ z ≤ L)
Hill, 2005, ch. 5.

( )APPENDIX

N1 ( z ) =A.1 Exact Solution of Pile Static Load Test [5]

(

)

( )

sinh α ( L − z ) 
sinh α (19)Assume that displacement at pile head in Fig.
1(a) is greater than or equals to yielding
displacement w * . Prescribe the boundary conditions(4) into the solution of governing Eqs. (3) and (5),obtain the solution of load at pile head as follows:=
P0 aa
EAw*  tanh (  + γ ) + a z0 

N2 ( z ) =sinh α ( z − z0 ) 
sinh α (20)

where L = z 0 +  is total length of pile element (Fig.
1(a)). Either elastic or plastic soil portions,subscripts “1” and “2” always refer to upper andlower nodes, respectively.

(14)

where the variable γ is the characteristic value of theend bearing stiffness of soil (k b ) at pile tip:516

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ARTIFICIAL SLUDGE BASED ON COMPOSITIONAL INFORMATION
OF A NATURAL SEA SLUDGE
Hirosuke Hirano1†, Takeshi Toyama2, Nobuyuki Nishimiya2, Davin H. E. Setiamarga 1,3, Shugo Morita1, Yuto
Uragaki1, and Kyoichi Okamoto2
1

National Institute of Technology, Wakayama College, Japan
College of Science and Engineering, Nihon University, Japan
3
Graduate School of Science, The University of Tokyo, Japan†: Corresponding author2

ABSTRACTThe chemical and physical properties of sludge collected from different aquatic areas differ from one another.
Such differences have made the universal standardization of early settings difficult. Usage of artificial sludge
with stable characteristics has been thus proposed in the present study. First, values of organic matter, sulfide
content, and the inorganic content were measured for the sea sludge collected from Funabashi Port in Japan.
Using the obtained chemical and physical properties as reference, we then produced artificial sludge by mixing
dry yeast made from yeast cultured in liquid sodium sulfide nonahydrate solutions (3.5-4.0%) and zeolite forinorganic material. In the present study, we understood that we couldn’t obtain the desired organic content for
the artificial sludge, simply by adding the yeast with the amount similar to the organic content percentage weaimed. We thus calibrated the amount of yeast by measuring the actual organic content of several artificial
sludge samples by thermo-gravimetry analysis. We found out that we need only to write a calibration curve of
three measurements (zeolite-dry yeast w/w% = 0%, mid-point, and 83.3%) in order to obtain the information
about the right amount of dry yeast addition to produce an artificial sludge with desired percentage of organic
material. Finally, we mixed 31.4% of dry yeast (zeolite and other inorganic materials = 68.6%), in order toobtain an artificial sludge with 23.6% organic content. This percentage is similar to the organic content of thesludge collected from Funabashi Port.
Keywords: Artificial sludge, Zeolite, Dry yeast, Water area

be affected by even the subtle differences ofenvironmental and weather changes, thus making
standardization possible. However, producing an
artificial sludge with characteristics mimicking thoseof the natural one is not easy, since many factors
must be calculated in during the production process.
Here, we report our result of producing an
artificial sludge useful for research requiring sludge
with its early values constant. Moreover, we alsoachieved our aim of constantly reproducing theartificial sludge, with stable standard values andeasy to reproduce.

1. INTRODUCTION
Sludge easily accumulates in closed water
system, such as lakes, ponds, and bays. This
problem occurs because such systems lack theability of extensive water circulation. The
accumulation of sludge might then cause problems
because the anaerobic condition of the sludge mightthen cause decompositions by microorganisms. In
order to solve this problem, it has conducted a studyof cleaning sludge in closed water area by using
aerobic microorganisms and micro bubbles [1]. Insuch studies, it is important to collect fresh sludge
from water area every time, because the organic
content of the sludge makes storage difficult.
However, collecting fresh samples every time makes
universal standardization of early settings difficult,because of differences in weather and environmental
conditions such as temperature and humidity.
In order to overcome the difficulties of setting
initial values, we propose the usage of artificial
sludge with stable characteristics for such studies.
An artificial sludge, with its early values set, will not

2. ANALYSES OF NATURAL SLUDGEPROPERTIESIn order to produce an artificial sludge useful forstudying the environment, we need the artificial
sludge to mimic the characteristic and properties of
the natural one. However, since natural sludge
samples taken from different locations and times
would show different starting values, it has needed
to choose one sample to use as the point of reference.
For this purpose, we decided to use the natural
517

GEOMATE- Bangkok, Nov. 14-16, 2016

samples taken from Funabashi Port in Chiba,
Northeast Japan, as the reference of the present
study. This is because Funabashi Port is locatedclose to our campus hence allowing multiple
sampling trips easy and practical, and since it has
been using the samples from the location in otherstudies (e.g. [2]).
Thus, in order to understand the naturalproperties of the sludge, in first studied material ofnatural sludge it has collected from Funabashi port
in Chiba, northeast Japan.

2.3 Sulfide
We then used gas detector tube to check thevalue of hydrogen sulfide (H 2 S). And about thevalue of H 2 S by gas detector tube, it was tried likethese conditions about sample water; by mixing 1 gof sludge and 100 cm3 of water. As a result, value
was 0.016 g/1g, so concentration of S in sludge was
1.5% by molecular weight.
3. ARTIFICIAL SLUDGE PRODUCTION
3.1 Preliminary artificial sludge production
In order to obtain the data about the amount ofmaterials needed to produce the final artificial
sludge, we conducted a preliminary experiment of
mixing materials to synthesize an artificial sludge. In
this experiment, it has mixed dry yeast and aqueousofsodiumsulfidenonahydrate
solution(Na 2 S∙9H 2 O), which was 3.5 to 4.0 mass%, thatratio was 1 g of dry yeast to10 cm3 of Na 2 S∙9H 2 O.
After discarding the supernatant, we collected the“dry yeast mud”, and mixed it with SiO 2 three times
the amount of the yeast mud. It was measured thistrial artificial sludge by TG and compared the value
with Funabashi’s sludge, but we were unable tomimic the organic content of the natural sludge, andthought that this was caused because the inorganic
content (sulfur compounds, etc.) in the yeast mud
was added to the total inorganic content of thepreliminary artificial sludge, causing a decrease in
the percentage of the organic content.

2.1 Organic contents
We conducted thermo-gravimetry (TG) analysis
in order to check the value of organic content. At
temperatures above 100ºC, moisture content of thesampled material will boil away, while above 600ºC,
the organic content of the material will completely
evaporate, leaving only the inorganic part.
Accordingly, in order to allow us to measure thepure value of the organic content, TG analyses wereconducted between the values of 100ºC and 600ºC.
The result of TG analyses on the natural sludge
sample collected from Funabashi Port showed thatthe organic content is 23.6% (Fig. 1, solid line). For
comparison, we also measured natural sludge
samples collected from other the Hidaka-port inWakayama in southwest Japan, where is locatedabout >350 km in Funabashi port in southwest
Japan. The Hidaka-port sludge has a 7.3% organic
content (Fig. 1, dashed line).
2.2 Inorganic contents
In order to check the property of inorganic matter,
we conducted energy dispersive X-ray spectrometry(EDX) on the samples. From EDX analyses, wefound out that this sample has the mineral content
of: silicon (Si), aluminum (Al), iron (Fe) and sulfur(S) (Table 1). X‐ ray diffraction (XRD) analysis
also showed that the Si content were silicon dioxide(SiO 2 ) and kyanite (Al 2 SiO 5 ) (Fig. 2).

Conditions of inorganic matter in sludge

Funabashi

Si [%]

1st
2nd
3rd
4th
5th
Hidaka
1st
2nd
3rd
4th
5th

0

Al [%]

Fe [%]

46.41

20.02

19.20

46.47

19.70

22.19

50.98

47.9

21.09

20.6

18.19

S [%]
14.37
11.64

21.3

9.74

46.17

21.19

24.69

7.95

49.26

20.88

22.12

7.74

Si [%]

Al [%]

Fe [%]

65.84

23.69

7.01

68.54

23.04

6.40

67.95

65.7

22.60

23.5

6.30

10.3

S [%]
3.47
2.02

7.4

3.16

62.65

24.27

8.69

4.39

63.42

23.92

8.68

3.97

3.4

-10
Intensity [-]Heat weight loss [%]Table 1

-20

-30
0

200

400

600

Temperature [ºC]
Fig. 1: Values of organic contents by TG○: in Hidaka port, ●: in Funabashi port
518

Angle of diffraction [2θ (CuKα)]Fig. 2 Components in a typical sludge detected by XRD○: SiO2, ●: Al2SiO5

GEOMATE- Bangkok, Nov. 14-16, 2016

We tried varying compositions of organic
contents (= yeast mud; 0%, 25%, 50%, 75%, and100% of yeast mud, of the total SiO 2 yeast mudmix). The results are plotted on the calibration
curves shown as Fig. 3 and Fig. 4. However, we still
could not mimic the value of the organic content of
Funabashi Port sludge when it has followed thevalues shown in the figures.
We then switched to use average of the values
in the first order and quadratic, and were successful
in obtaining an artificial sludge with the organic
content of 23.8%, which was within 1.0% error. In
second trial, it was successful to make 23.6% oforganic matter in artificial sludge. Although reasons
why we were able to reproduce the value of organic
matter in artificial sludge by averaging the first orderand quadratic values was still unknown, it was able
to set the value of organic content in artificial sludge
by any numerical values, if we use the averagevalues of the callibration curve mentioned above.
We also found an interesting characteristic of thecalibration curves in Fig. 5: calibration curves byapproximation formula of first order and quadratic
must intersect at 83.3%. It is unknown about thisreason too. However, this information helps indrawing an easier calibration curve: rather thanwriting the intersection points at several standard
points, it is able to use the calibration curves 0% and83.3%, and the middle point.

Measured valuesof organic matter [%]Measured values of S [%]3.2 Zeolite addition
After success on finding the right amount oforganic matter addition as explained in the previous
section, it was tried to use artificial sludge toreproduce previous studies about sea sludge, and see
if the results of such studies were reproduced, whenusing artificial sludge. We chose trying to reproduce
previous experiments about cesium ion (Cs+)
decontamination [3], and set the experimental
conditions to be similar with that. For the study, it
was added cesium nitrate (CsNO 3 ; 1000 ppm) topreliminary artificial sludge (dry yeast and SiO 2 ;totaling to 10 g). We prepare 5 samples of the mix,
and then further added 20cm3 of hydrogen peroxide
solution (H 2 O 2 ) with 5 types of concentrations,
between from 0% to 34.5%.
In the present study, no decontamination effectwas observable in artificial sludge like what we
expected (Fig. 6: white dots = expected result, black
dots = obtained result). This suggested that theartificial sludge could not adsorb Cs+, and it wasdifficult to show the effect of decontamination
reproducively as in the previous study [3], by the

Adding values of organic matter [%]Fig. 5 Calibration curve in artificial sludge

Measured valuesof organic matter [%]Measured values of Cs+[Measured values / First value]

Adding values of organic matter [%]Fig. 3 Calibration curve in trial (1)Adding values of organic matter [%]Fig. 4 Calibration curve in trial (2)Added values of CsNO3 [ppm]Fig. 6 Results of Cs+ decontamination○: Funabashi’s sludge, ●: Artificial sludge by SiO2
519

GEOMATE- Bangkok, Nov. 14-16, 2016

lack of as the sludge, or in other words, the absenceof adsorbed Cs+ in the first place.
Previous studies have shown that in sludge, theorganic content has Cs+ adsorption ability (e.g. [4]and [5]). However, our study here also suggests thatbesides the organic content, the inorganic content
has a Cs+ adsorption ability (Table 2) by [6], which
was lacking in preliminary sludge.
Based on the result of XRD analysis, we added
zeolite (vermiculite, kind of silicate mineral) to
preliminary artificial sludge for the inorganic
material, in order to add the Cs+ adsorption ability ofartificial sludge. Another reason of choosing zeolite
as one of the inorganic content ingredients was
because previous studies have shown that zeolite has
the ability of Cs+ adsorption (e.g. [6]), similar to the
inorganic matter in natural sludge, which is also a Si
compound (e.g. kyanite).

expected values (22.6% and 23.0%, respectively).
4. CONCLUDING REMARKS
4.1 An artificial sludge as a model system forenvironmental studies
In previous studies, we have analyzed thecharacteristics of sludge such as cesium adsorption[4] and [6], and cesium decontamination by [3]. We
used the obtained data from such analyses asreferences when making artificial sludge.
In the present study, in order to obtainpreliminary data to produce an artificial sludge withthe characteristics mimicking those of the naturalone, we initially studied material of natural sludge
collected from Funabashi port. We also produced
different preliminary artificial sludge, by trying
various amount of compositions and materials.
These preliminary data have allowed us to find thebest composition and materials to utilize in order toproduce artificial sludge.
We used dry yeast in order to recreate theorganic content of artificial sludge, because it has
similar ability of Cs+ adsorption as the organic
matter collected from the supernatant of a sludge
suspension liquid (Fig. 3). Moreover, that will belike mud when added water. Also, a natural sludge
obtains its organic content from biological sources
such as carcasses of living organisms. Using dry
yeast as the organic content source will mimic thiscondition. Dry yeast is also easy to obtain, making
artificial sludge production easy.
We used SiO 2 and zeolite in artificial sludge forthe inorganic material, basing on the result of XRD
analysis. Moreover, previous studies have shownthat zeolite has the ability of Cs+ adsorption (e.g.
[5]), similar to the inorganic matter in natural sludge,
which is also a Si compound (e.g. kyanite).
EDX analyses showed that the organic content
of a natural sludge also contain sulfur. In order tomimic such condition, it was added Na 2 S∙9H 2 O to
dry yeast. The amount of the added Na 2 S∙9H 2 O wasfound to be 3.5 to 4.0 mass% in aqueous solution of
Na 2 S∙9H 2 O for the dry yeast culture (Fig. 7).

3.3 Final result
After succeeded in deciding the best compositionof materials for artificial sludge through various
trials and analyses, we synthesized final artificial
sludge, and then analyzed its characteristics andvalues. We plot the results in another calibration
curves as shown in Fig. 5. According to the new
calibration curves, it was found that in order toreproduce Funabashi sludge, we found to need the
addition of 28.8% organic materials to the total
mixture. To test the accuracy of recipe based on new
calibration curves, and made additional artificial
sludge with various amount of organic material
additions (27.1% and 27.7%). The resulting organic
contents based on TG analyses were 23.2% and22.6%, respectively. These values theoretically areless than 1% of error, when compared to theTable 2

Adsorption ability of zeolite

Nothing to add

640.7

-

Zeolite (Large)

632.4

8.2

Sea sand

637.3

3.4

SiO2

641.2

-0.5

Inorganic matter bysludge

621.4

19.3

Measured values of S [%]First value of Cs +Captured [ppm/g][ppm]

Captured values by
Values of Cs + Values of Cs +(Before) [ppm] (After) [ppm] inorganic matter [ppm/g]
Inorganic matterin sludge （1 g）

640.5

Large
Zeolite
（1 g）

Small

647.5

PowderAvarage

-

588.9

51.6

617.7

29.8

607.8

39.6

536.4

111.1

-

60.2

Fig. 7520

Na2S∙9H2O [mass%]
The best value of adding Na2S∙9H2O

GEOMATE- Bangkok, Nov. 14-16, 2016[2] Hirano H, Nishimiya N, Kojima Y, Toyama T,
Umegaki T, Okamoto K, “Investigation ofradioactive cesium decontamination from sea
sludgeusing
microbial
activity”,
The
International Symposium on Inorganic andEnvironmental Materials 2013, 2013.
[3] Hirano H, Toyama T, Nishimiya N, Okamoto K,“Removal of Cesium from Sea Sludge throughDecomposition of Organic Matter with Aqueous
Hydrogen Peroxide”, International Journal of
GEOMATE, Vol. 8, No. 1, 2015, pp. 1203-1206[4] Uragaki Y, Morita S, Hirano H, “Ability ofadsorbed cesium by organic matter from sea
sludge, by Name of Presenter”, 50th Annual
Conference of Japan Society on Water
Environment, 2016, pp. 622[5] Nakao A, “Cesium Adsorption and Fixation on
Soils”, TREND IN THE SCIENCE, Vol. 17, No.
10, 2012, pp. 10_40-10_45[6] Morita S, Uragaki Y, Hirano H, “Comparison ofadsorbed ability for cesium, by inorganic matterand sea sludge”, 50th Annual Conference ofJapan Society on Water Environment, 2016, pp.
621[7] Hirano H, Toyama T, Nishimiya N, “Preparationof prototype artificial sludge based oncompositional information of sea sludge”, 48th
Annual Conference of Japan Society on Water
Environment, 2014, pp. 298
Hirano H, Uragaki Y, Morita S, “Preparation of
prototypeartificial
sludgebasedoncompositional information of sea sludge Part 2:
Improvement idea about contents”, 50th Annual
Conference of Japan Society on Water
Environment, 2016, pp. 190

4.2 Future prospects
In the present study, we had tried to use artificial
sludge for experiment of Cs+ decontamination. Thus,
we can safely say that artificial sludge can bebeneficial for future experiments about Cs+
decontamination.
About results with H 2 O 2 addition to artificial
sludge containing CsNO 3 , Cs+ was leached out by
the H 2 O 2 into the liquid phase. This has caused themeasured values to differ from result by Fig. 6. But
when we compared this result and a previous study
using the natural sludge, our artificial sludge did not
reproduce the study. By the way, natural sludge
samples collected from Funabashi Port and Hidaka
Port, showed similar values, suggesting that thedecontamination ability of these natural sludge
samples taken from two different areas could be
similar [5].
Future studies would also include trying to
reproduce the characteristics of the natural sludge
taken from various locations with a variability ofdiffering characteristics, conditions, and setting
values. Such studies will allow for further testing of
the robustness of our result reported here.
5. REFERENCES[1] Hirano H, Nishimiya N, Kojima Y, Toyama T,
Umegaki T, Hotta K, Okamoto K, “Purification
experiments on sludge by micro–bubble andmicroorganisms activator – using detergent
including enzyme as microorganisms activator –”, Proceedings of the 23rd Ocean Engineering
Symposium, 2012.

521

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

COLONIZATION AND MORPHOLOGICAL CHANGES OF A SEDGERESTRICTING REGENERATION AFTER WIND DAMAGE IN A
NATURAL FOREST IN KISO DISTRICT, CENTRAL JAPANTeruo Arase, Tetsuo Okano and Tetsuoh Shirota
Faculty of Agriculture, Shinshu University, Japan
ABSTRACT
Many old trees were blown down by typhoons in 1959 and 1961 in a natural coniferous forest deep in themountains of Kiso District, Central Japan, and dense dwarf bamboo subsequently delayed the growth of tree
seedlings. Forest engineers have tried to suppress dwarf bamboo to promote the regeneration of arboreal
vegetation. However, after suppression of dwarf bamboo, an unknown grass unexpectedly colonized the area andreplaced them. Afterward, almost no tree seedlings could be observed in the newly formed community, which
seemed to worsen the extent of regeneration. It has been identified that the grass as Carex oxyandra, a nativesedge species indigenous to Kiso District. Commonly, Carex oxyandra grows as short as about 10 cm like lawn
grass, but it seems to have exceedingly enlarged the size in the community. Thus, investigation on themorphological variation of this sedge in Miure Experimental Forest within the Kiso National Forest was carriedout. Leaf blade length, basal tiller length, and number of leaves per tiller were measured in upper and lower
stands on three slopes in 2009. Results revealed that leaf blade length and basal tiller length in lower stands wereapproximately twice as large as those in upper stands on each slope, whereas the number of leaves per tiller was
almost the same (8.8-9.4 leaves). Consequently, tussocks of this sedge became large in lower stands on a slope,which made the community overcrowded and damp, restricting tree regeneration.
Keywords: Natural Forest, Regeneration, Dwarf Bamboo, Carex oxyandra, Cilonization
famous timber brand, ‘Kiso-Hinoki’. The trees havebeen selectively cut for use as high-quality timber,and the trees have naturally regenerated for several
hundreds of years. However, many old trees
including Japanese cypress were blown down by
typhoons in 1959 and 1961. Subsequently, dwarf
bamboo (Sasa sp.), the dominant species on theforest floor, remained dense on the land surface [3],[4].
Forest engineers and researchers have tried topromote the regeneration of arboreal vegetation.
They deduced that the seedlings had not successfully
established, and suggested that the dwarf bamboo
should be suppressed. The growth of Japanesecypress seedlings requires a very limited
environment: it demands a shaded habitat with arelative luminous intensity from 2 to 5%, andmoderately humid soil. However, the relative
luminous intensity reaches only 1 to 2% under adense community of dwarf bamboo. In addition, thecool pluvial climate in this district (an averageannual temperature of 7 C and an annual
precipitation of 3500 mm at the meteorologicalstation, which is at an elevation of 1300 m [4]) has
caused formation of a wet podzol layer in the soil,
which is too humid and nutritionally poor for theseedlings. Thus, after the wind damage from thetyphoons, the dense community of dwarf bamboo
prevented the growth of tree seedlings in this district,

INTRODUCTION
After weather damage in natural forests far fromhuman habitats, natural regeneration of trees is
expected, instead of requiring artificial introductionof vegetation by revegetation. If vegetation damage
occurs in a natural forest where germination andestablishment of tree seedlings is difficult for a long
period, technology promoting the growth of
spontaneous seedlings is needed.
Recently, revegetation technologies have beendeveloped and applied especially to cut slopes andlandslide slopes adjacent to human habitats. This
involves construction of foundations for retaining
the soil and introducing vegetation. It is important tomonitor succession of vegetation after its
introduction, since plants will grow or decline, andthe species composition is altered by surrounding
species year after year. As an example, it wasreported that the process of vegetational succession
on landslide slopes on Mikura-jima Island, where
native tree and herb species were initially plantedusing simple terracing [1], [2]. In contrast,
regeneration technology in natural forests has
scarcely been studied in Japan.
In a natural coniferous forest deep in themountains of Kiso District, Central Japan (Fig. 1),
large Japanese cypress (Chamaecyparis obtusa)
trees aged up to about 300 years are known by a522

GEOMATE- Bangkok, Nov. 14-16, 2016

dwarf bamboo (Fig. 2). Afterward, few tree
seedlings could be observed in the newly formed
community, which seemed to reduce regeneration
further.
Forest engineers at first suspected that the grassmight be an invasive foreign species, but it has been
identified as Carex oxyandra, a native sedge species
indigenous to Kiso District. Commonly, Carex
oxyandra grows as tussocks as short as about 10 cm
like lawn grass, but it seems to have exceedingly
enlarged the size in the community. Carex oxyandra
is reported to remain at the forest floor as a lowpreference foliage by sika deer [5], and to colonize
bare ground after mining as a pioneer plant [6], butcases of colonizing in a forest or grassland where
vegetation has already developed were previously
unknown.
In the present study, the investigation of themorphological variation of Carex oxyandra in the
Miure Experimental Forest within Kiso National
Forest was carried out to identify the replacement
dwarf bamboo by this sedge.

Fig. 1 Location of study site

METHODS
The study site was located in Miure
Experimental Forest, in the center of the mountain in
Kiso District, at an elevation from 1400 to 1500 m.
In October 2009, we observed communities ofCarex oxyandra on three slopes (forest
compartments No. 2626, 2627 and 2630), andselected both upper and middle stands on each slope.
Leaf blade length, basal tiller length (reddish-purple
part near the base) and number of leaf blades pertiller were measured for five individuals per strand.
The environmental conditions were as follows:(a) Stands in No. 2626 were located at an elevationfrom 1400 to 1420 m, with a slope direction of 0
and an inclination of 27 . Dwarf bamboo were
withered (dry culms remained), and the communityof the sedge was observed on bare ground.
(b) Stands in No. 2627 were located at an elevationfrom 1500 to 1520 m, with a slope direction of 260
and an inclination of 30 . Communities wereobserved where dwarf bamboo dominated and wheresedge dominated.
(c) Stands in No. 2627 were located at an elevationfrom 1480 to 1500 m, with a slope direction of 220
and an inclination of 20 . Dwarf bamboo was rarely
observed, and the sedge constituted the largest
community.
In each stand, arboreal vegetation (comprised ofJapanese cypress and some broad-leaved deciduoustrees) was sparse, and plants with sizes common for
Carex oxyandra were observed in the area near the
ridge above the upper stand. The difference inelevation between upper and middle stands on eachslope was approximately 10 m.

Fig. 2 A case of colonizing Carex oxyandra(at forest compartment No. 2630) after
withering of dwarf bamboo
especially in the higher elevation areas [3], [4].
In 1966, the Miure Experimental Forest of the
Nagano Regional Forest Office was established tofind a suitable regeneration method for forests in thewet podzolic zone, which appears in the peneplainarea (at an elevation of about 1200 to 1500 m) on thesouthwest slope of Mt. Ontake in the Kiso District[4]. Many forest compartments (divided sections forforest managements or experiments) were
established to examine various methods ofregeneration involving the suppression of dwarf
bamboo. In successful forest compartments, dwarf
bamboo was suppressed by herbicide for several
years, and the seeds of Japanese cypress germinatedduring this period and grew in moderate shade under
the community of regenerating young dwarf bamboo.
However, in 2006, an unexpected phenomenon
was observed in higher-elevation areas: an unknown
grass began to colonize and replace the withered
523

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 4

Herbarium specimens of Carex oxyandra

collected on the slope No. 2630 (left: middle
stand; right: upper stand). The black scales in the
figure are 16 cm.
significant differences based on specific slopes weredetected (F-test).

Fig. 3 The leaf blade length and basal tiller
length of Carex oxyandra in each stand.
Solid columns (left) and hatched columns(right) show the middle stand and the upper
stand on the slope, respectively. The bar on
the top of each column indicates ±1 SD.

In contrast, the number of leaf blades per tiller
proved relatively stable among stands; the average
was 8.2 to 9.0 in upper stands, and 8.8 to 9.4 inmiddle stands. No significant differences weredetected based on either location on a slope or on
specific slopes (F-test). In Fig. 4, we present imagesof actual samples of the sedge on the same slope (No.
2630) that were obviously different from each other:
the number of leaves per tiller did not seem to differ,
but the leaf blade length and basal tiller length in the
middle stand were twice as large as in the upper
stand.
DISCUSSIONRESULTS
Fig. 3 shows the leaf blade length and basal tiller
length in each stand. The average leaf blade length
was 7 to 15 cm in upper stands, whereas it was 20 to
30 cm in middle stands. The maximum length was
36.0 cm (in the middle stand of No. 2626 slope).
Analysis of variance detected a significant
difference based on location on a slope (F-test,
p<0.01). A significant difference based on specificslope was also detected (F-test, p<0.05), since thelengths were smaller on No. 2627 slope than theothers.
Average basal tiller length was 1.0 to 2.2 cm inupper stands, whereas it was 3.3 to 3.6 cm in middle
stands (Fig. 4). The maximum length was 5.2 cm (inthe middle stand of No. 2627 slope). Analysis ofvariance detected a significant difference based on
location on a slope (F-test, p<0.01). However, noAt the study site, the native sedge Carex
oxyandra proved to change the above-groundmorphology drastically. Between communities close
to each other on the same slope, leaf blade lengthand basal tiller length in middle stands were twice aslarge as in upper stands (Figs. 3 and 4). In addition,
the number of leaf blades per tiller was relatively
stable independent of the location on a slope,
meaning the sedge community was dense in the
middle stands.
For Carex oxyandra, the upper and middle stands
in each slope were not considered to be genetically
remote, since the distance between them was
relatively close. The morphological changes in thissedge were presumably the result of adaptation to
growth environments, such as soil moisture and airtemperature. For example, water stress and highte mp er atur e ar e r epo r ted to infl uence lea f
524

GEOMATE- Bangkok, Nov. 14-16, 2016

grasses on the fascined landslide slopes in
Mikura-jima Island, Japan”, International
Journal of GEOMATE, Vol. 8 (2) 2015, pp.
1316-1322[2] Okano, T. and Arase, T., “Vegetation recovery
process on landslide steep slope after Alnus
sieboldiana and Miscanthus condensatus
planting with simple terracing work in Mikurajima Island, Japan”, International Journal of
GEOMATE, Vol. 10 (3) 2016, pp. 1884-1890[3] Akai, T. and Asada, S., “Studies on Natural
Regeneration (I) The reproduction of
Chamaecyparis at wet podozolic zone in Kiso
District”, Bulletin of the Kyoto University
Forests, Vol. 39, 1967, pp. 35-63. (in Japanesewith English Résumé)[4] Akai, T., “Studies on Natural Regeneration (II)The regeneration of Chamaecyparios obtusa at
Miure Experimental Forest in Kiso District”,
Bulletin of the Kyoto University Forests, Vol.
44, 1972, pp. 68-87. (in Japanese with English
Résumé)[5] Yoshikawa, M., Imafuku, H. and Hoshino, Y.,“Distribution of forest floor vegetation after thedisappearing of dwarf bamboo in thesenjugahara area, Oku-Nikko”, Journal of theJapanese society of Revegetation Technology,
Vol. 39, 2014, pp. 368-373 (in Japanese withEnglish abstruct)[6] Tsujimura, A., “The ecology of Carex oxyandra
II. The behavior of seedling and tillers”,
Ecological Research, Vol. 2 (3), 1987, pp. 279288.
[7] Yoshikawa, K. and Wang, L.H., “Effects ofwater stress on leaf developments of several tree
species”, Journal of the Japanese Society ofRevegetation Technology, Vol 16 (4), 1991, pp.
18-30 (in Japanese with English summary)[8] Sasaki, O., “Effect of high water temperature on
the morphology of crown roots and leaf in rice
plants”, Japanese Journal of Crop Science, Vol.
61 (3), 1992, pp. 388-393 (in Japanese withEnglish abstruct)[9] Tanaka, K., Yamawo, A. and Yano, O., “Seeddispersal by Ants in Carex oxyandra var.
oxyandra (Cyperaceae) from Japan”, The
Journal of Japanese Botany, Vol. 90(2), 2015,
pp.129-133

morphology in some grasses [7] and trees [8]. If
excessive soil moisture favors large tillers and leaf
blades of Carex oxyandra, the above-groundenvironment would also become wetter, which
impedes the establishment of Japanese cypress

Fig. 5 Schematic depiction of the presumed
process of Carex oxyandra colonization after
withering of dwarf bamboo
seedlings. Therefore, a method to suppress thissedge is needed for the regeneration of Japanesecypress at the study site.
Finally, we try to explain why Carex oxyandra
replaced dwarf bamboo. There are two means of
expansion, vegetative reproduction (clonal growth)and seed dispersal. From the viewpoint of vegetative
reproduction, tillering from the base is one means,since this sedge produces many tillers.
However, the sedge is able to expand only
incrementally outward merely by clonal growth.
From the other viewpoint of seed dispersal, it has
been reported that Carex oxyandra seeds are
dispersed by ants: an elaiosome is attached to theseed surface, and ants take the seeds to their nest andbite off only the elaiosome [9]. Consequently, wesuggest a process for colonizing by Carex oxyandra
in Fig. 5. How long the dominance of the sedge lasts
or whether dwarf bamboo can recover in the sedgedominated community are worth investigating interms of the regeneration of the forest.
REFERENCES[1] Arase, T. and Okano, T., “Influence offertilization on native plants and exotic pasture525

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

METHODS OF SUPPRESSING COLONIZING SEDGE TO HELPTO ESTABLISH TREE SEEDLINGS IN A NATURAL FOREST
IN KISO DISTRICT, CENTRAL JAPANTeruo Arase, Tetsuo Okano and Tetsuoh Shirota
Faculty of Agriculture, Shinshu University, Japan
ABSTRACT
Dense bamboo grasses have delayed the growth of tree seedlings after the wind damage in a naturalconiferous forest in deep mountains of Kiso District, Central Japan. After suppression treatment of the bamboo
grass, a native sedge Carex oxyandra replaced them and made the growth of tree seedlings worse. The techniqueto suppress colonized Carex oxyandra is unknown, since this sedge has not ever been a harmful weed to becontrolled. Thus, we established first experimental plots for suppressing Carex oxyandra by a herbicide,
glyphosate potassium salt solution (Roundup Max load) by normal levels of dilution (control without herbicide,
diluted to ×25, ×50 and ×100 volume with water) in May 2010. Next experimental plots with above-normal
levels (diluted to ×100, ×200, ×400, ×800, ×1600 and ×3200) were established in June 2011. Coverage
percentage, plant height of Carex oxyandra and tree seedlings were measured in each plot in autumn up to 2015.
As a result, coverage and plant height of Carex oxyandra decreased obviously after a half year in all dilution
levels, and the suppressing effect continued during 2 years in the lower dilution from ×25 to ×400. However,withered leaves formed a carpet-like thick mat, which remained several years and was suspected to restrict thegermination and growth of tree seedlings. Consequently, the herbicide proved out effective to the sedge for 2
years, but some additional treatment is needed to secure the foundation space for tree seedlings within the mat ofwithered leaves.
Keywords: Carex oxyandra, herbicide, Japanese cypress, seedling, natural forest
sliding surface and induces gliding of the snow
cover and avalanches [5]. On the other hand, thedwarf bamboo community plays a significant role inretention of basic cations in the surface soil andprevention of soil acidification [6].
There are several methods for suppressing dwarf
bamboo, such as weeding, application of herbicide,and grazing by herbivorous livestock [7].
Application of herbicide, i.e., chemical control, hasbeen preferred because its effects last a long time.
Forest engineers and researchers have suppresseddwarf bamboo by herbicide application since 1967
in Miure Experimental Forest in the Kiso District topromote regeneration of the forest. However,
unexpected colonization by Carex oxyandra, anative sedge indigenous to this area, was found after
the withering of dwarf bamboo following herbicide
treatment in higher-elevation areas in 2006.
Afterward, almost no tree seedlings could beobserved in the newly formed community, which
seemed to indicate even poorer regeneration [8].
It has been reported that exposure to light
improves germination of Japanese cypress seeds,
with a typical germination percentage from 10 to
40%, which varies according to area and year [9].
However, the growth of seedlings requires a shaded
habitat with a relative luminous intensity from 2 to
5% (germination is suspected to be impaired at a

INTRODUCTION
In a natural coniferous forest deep in themountains of Kiso District, central Japan, many old
trees including Japanese cypress (Chamaecyparis
obtusa, known as the famous timber brand ‘KisoHinoki’) were blown down by typhoons in 1959 and1961. Subsequently, dwarf bamboo (Sasa sp.), thedominant species in the forest floor, remained denseon the land surface. In addition, the wet podzolic soil
formed by the cool pluvial climate in this district (anaverage annual temperature of 7 C and an annual
precipitation of 3500 mm at the closest
meteorological station at an elevation of 1300 m)
was too humid and nutritionally poor for seedlingestablishment. Thus, after the wind damage due tothe typhoons, the regeneration of trees in this district,
especially in the higher-elevation areas, wasprevented [1], [2].
In forestry of afforested land, suppressing forestfloor vegetation (dwarf bamboo or other shrubs inJapan) is important [3]. Weeds influence planted
trees by interception of light and capture of nutrients
and water [3]. In dwarf bamboo, suppression of thecommunity is reported to enhance the species
diversity of forest floor vegetation and the density ofregenerated trees [4]. If the dwarf bamboo
community falls under snow cover, it acts as a
526

GEOMATE- Bangkok, Nov. 14-16, 2016

relative luminous intensity over 10%) [1], [2].
Japanese cypress seeds are reported to lose
germination ability after 1 year in field conditions[9], and some seeds suffer predation by animals [10].
Therefore, the seeds must germinate rapidly in theabsence of covering vegetation, and then theseedlings need to be shaded by recovered vegetation.
The effect of herbicide suppressing dwarf bamboo isestimated to last about 3 years based on remotesensing [7], which is reasonable for the regenerationof Japanese cypress trees: seeds can germinate in abright environment and then the seedlings can growwhile shaded by recovered dwarf bamboo.
However, no method is yet known to suppress
Carex oxyandra, because harmful colonizing by thissedge is uncommon. Considering the cost and labor
for forest management, it is reasonable to employ amethod similar to that of suppressing dwarf bamboo
as a first approach.
In the present study, we examined the effect ofherbicide at various dilution levels on suppressing
the community of Carex oxyandra in the Miure
Experimental Forest within the Kiso National Forest.
The influence of suppressing the sedge on theestablishment of Japanese cypress seedlings is also
discussed.

METHODS
Study Site
The study site was located in forest compartment
No. 2630 in the Miure Experimental Forest, thedeepest area in the Kiso District, at an elevation of1500 m. At the site, most of the dwarf bamboo had
already withered or disappeared, and the sedge
formed a large-scale community on an almost level
land surface (slightly inclined to the north).
Experiment 1:
Herbicide at Normal Levels of Dilution
In May 2010, we applied a generally circulated
herbicide, glyphosate potassium salt solution (brand
name: Roundup Max load), to the sedge community
at four levels of dilution (control without herbicide,and herbicide diluted with water to ×25, ×50 and×100 by volume).
For each of the four treatments, an application
zone of 15 m2 (3 m ×5 m) involving three fixedsurvey 1-m2 plots (1 m × 1 m) were designed, andtwo blocks were established. The experimental
design was a randomized block method with oneway layout (4 dilution levels ×2 blocks = 8
application zones).
After application of herbicide, seeds of Japanesecypress were sown that had been harvested fromsurrounding trees in 2009. Sowing density was 6,000
seeds per zone (i.e., 400 seeds per m2).
Coverage percentage and height of Carex
oxyandra were measured yearly in fall (from midOctober to early November), from 2010 to 2015. In
measuring the plant height, we measured 5 plants
per plot randomly (or all plants if the number ofsurviving plants was less than 5) and averaged thevalues. We also observed and measured thethickness of the carpet-like mat of withered leaves
that formed after the application of herbicide (Fig. 1).
On the same day, the number of seedlings and heightof each Japanese cypress seedling were measured.
Experiment 2:
Herbicide at Above-normal Levels of Dilution
In June 2011, we applied the same herbicide to
the sedge community at six more dilute levels(diluted with water at ×100, ×200, ×400, ×800,
×1600 and ×3200 by volume).
For each of the six levels, an application zone of
4 m2 (2 m ×2 m) was established. Only one plot wasestablished for each level, because we intended thisexperiment as a preliminary one to determine
whether above-normal levels of dilution (i.e. lower
concentrations of herbicide) would be effective. The
plots were established in a community of Carex
oxyandra near the experimental zones treated at

Fig. 1 Suppression of Carex oxyandra community
by herbicide. Top: inside (left) and outside thezone of herbicide application (right). Bottom:
carpetlike thick mat of withered leaves. Bothphotographed in fall 2010, half a year after
herbicide application.
527

GEOMATE- Bangkok, Nov. 14-16, 2016

became more obscure year by year, and in 2015
reached 60 to 90% of control coverage, and theheight of plants reached 11 to 17 cm. However, theextent of sedge coverage had unexpectedly declined
on a massive scale in all plots, involving even thecontrol, in 2015.
Table 1 demonstrates the growth of Japanesecypress seedlings in the experimental plots. In
control plots, no seedlings were observed till 2015.
In other plots, the lower the concentration ofherbicide, the more germinated seedlings were foundin fall 2010 (from 7.4 to 22.2 individuals per m2).
However, in 2011, the number of seedlings
decreased markedly in each plot (to 1.5 to 3.2
individuals per m2). In 2012, the number of
seedlings also decreased (to 1.0 to 1.9 individuals
per m2), whereas the plants generally increased byapproximately 2 cm in height. A gradual decrease in
number of individuals and plant height occurred
subsequently. Finally, in 2015, seedlings survived
only in the plot treated at an herbicide dilution levelof ×100 (0.4 individuals per m2), in which someseedlings elongated by over 10 cm in height.
The thickness of the mat of withered Carex
oxyandra leaves was 1.5-2.5 cm in each stand in fall2010 (Fig. 1), half a year after herbicide application

normal dilution levels; the elevation, inclination andlandform were similar to them.
Coverage percentage and height of Carex
oxyandra were measured yearly in fall (from midOctober to early November), from 2010 to 2015.
RESULTS
Experiment 1:
Herbicide at Normal Levels of Dilution
Changes in Carex oxyandra growth after
treatment are shown in Fig. 2. In fall 2010, thecoverage percentage of the sedge decreased to lessthan 5%, and plant height also decreased by over 10
cm at dilution levels of ×25, ×50 and ×100. The
differences from the control in coverage percentage
and plant height were significant in fall 2010 and2011 (Dunnett’s test, p<0.05). However, after 2012,
the sedge recovered and the differences from control

Table 1 Mean number of Japanese cypress seedlings
per m2 after applying herbicide at normal levels of
dilution to a Carex oxyandra community. Sowing
density of Japanese cypress seeds was 400 seeds
per m2.
LevelsYear

Ranks of plant height (cm)0-control

×100

2-

0.0
0.0

2013

0.0

2014
2015

0.0
0.0

2010
2011

8.0

2012

0.2

13.8
2.2

0.3
0.8

0.3

2010
2011

0.3

0.3

0.2

0.2

0.2

0.2

0.3

0.2
0.7

5.3

0.2

0.2

0.7

0.2

0.2

13.2
3.2

0.2
0.3

0.8

0.2

7.8
1.2
0.7

2010

2.0

5.2

0.2

0.5

0.8
0.3

0.2
0.5

2014
2015

0.2

0.2
0.2

2011
2012
2013

22.2
3.2

0.2

2014
2015

528

Total
100.0

2012
2013

Fig. 2 Changes in Carex oxyandra growth
after applying herbicide at normal levels of
dilution

8-

2011
2012

2015

×25

6-

2010

2013
2014
×50

4-

0.2

1.3

1.8
0.5
0.2
0.0
7.3

0.2

1.5
1.2
0.2
0.0
0.0

GEOMATE- Bangkok, Nov. 14-16, 2016

and negatively correlated with the level of dilutionof herbicide in 2012 (Spearman’s rank correlation, rs=-0.83, p<0.05), but no significant correlation was
detected in 2013.
In contrast, plant height showed a complicatedfluctuating pattern. During 2011 and 2012, it slightly
increased or leveled off in the plots treated at ×100,
×200 and ×400, but decreased or leveled off in theplots at ×800, ×1600 and ×3200. The plant heightincreased in all plots during 2012 and 2013, butshowed diverse changes in 2014. In 2015, the plantheight had also unexpectedly declined on a massive
scale in most plots. Plant height was significantly
and negatively correlated with the level of dilutionof herbicide in 2013 (Spearman’s rank correlation, rs=-0.93, p<0.05), but no significant correlation was
detected in 2014.

at dilutions of ×25, ×50 and ×100. The thickness
decreased to 0.5-1.0 cm in 2011, 0.3-1.0 cm in 2012,
and 0-0.5 cm in 2013. In 2014, the mat seemed tohave decomposed and had disappeared from all
stands.
Experiment 2:
Herbicide at Above-normal Levels of Dilution
Changes in Carex oxyandra growth after
herbicide treatment are shown in Fig. 3. In fall 2011,
the coverage percentage of the sedge drasticallydecreased to less than 5% even in the plot at thehighest dilution of ×3200. The coverage percentage
slightly increased in 2012 in the plot treated at lower
levels of concentration, and obviously recovered ingeneral in 2013 (25 to 90%) and 2014 (50 to 90%).
However, the coverage percentage suddenly
decreased in 2015 (to 1 to 50%) due to unknown
causes. The coverage percentage was significantly

DISCUSSION
Suppression of Carex oxyandra
In experiment 1, herbicide diluted to normal
levels suppressed communities of Carex oxyandra infall 2010 and 2011, i.e., for 2 years after treatment.
After 3 years, the sedge steadily recovered andrecolonized the area (Fig. 2).
Also in experiment 2, herbicide diluted to abovenormal levels suppressed communities of the sedge
in fall 2011 and for the whole of 2012, i.e., for 2
years after treatment. After 3 years, the sedge
steadily recovered and recolonized the area (Fig. 3).
In both experiments 1 and 2, the plant height wassignificantly decreased by the herbicide for a limited
period. However, the plant height seemed tofluctuate, which is probably attributable to theaveraging process used in the study: if some slightly
damaged large plant survived among small plantsthat regenerated after withering, the average plantheight would be a middle value between them.
In a similar case of herbicide application to sedge
in a natural field, it has been reported that Carex
kobomugi, an invasive foreign species in coastal
dunes, was reduced but not eliminated after repeatedherbicide (Roundup) application [11]. In comparison,
Carex oxyandra was suppressed for approximately
two years by herbicide treatment, so careful choiceof herbicide and examination of its effectiveness areneeded according to the target plant species.
However, the reason for the sudden decrease inCarex oxyandra in 2015, which simultaneously
occurred in both experiments 1 and 2, was not
explainable in the present study. It might have beencaused by some extraordinary meteorological event,such as an extremely heavy snow or a long period ofwhen snow remained that is unusual in this area,
because this sedge is an evergreen species. It mightalso have been caused by the innate life span of thecommunity: some sedge species have been reported

Fig. 3 Changes in Carex oxyandra growth
after applying herbicide at above-normal
levels of dilution
529

GEOMATE- Bangkok, Nov. 14-16, 2016

levels, and the suppressing effect continued during 2
years in the lower dilution from ×25 to ×400.
However, the influence of suppressing the sedge on
the establishment of Japanese cypress seedlings isalso confirmed: withered leaves formed a carpet-likethick mat, which remained several years and wassuspected to restrict the germination and growth oftree seedlings.
Our results suggest this method to promote theestablishment of trees if an established communityof Carex oxyandra can be suppressed:(a) Scratching treatment (to breaking up mats)should be done after the community of Carex
oxyandra has been suppressed. Based on ourobservations, a small cutting tool like a hand dredge
or sickle is sufficient to break up or remove the mat.
Though the mat is thick just after withering(maximum thickness was 2.5 cm in the present
study), it is not so hard as to be difficult to break,unlike dwarf bamboo.
(b) If the establishment of tree seedlings is not
sufficient, herbicide should be reapplied to preventthe sedge from recolonizing.

to decline spontaneously after forming dominantcommunities, without any treatment to suppress
them [12]. Monitoring Carex oxyandra communities
is also desirable, since it not known whether thecommunity is transient or more permanent.
Establishment of Japanese Cypress Seedlings
Seedlings of Japanese cypress were notsuccessfully established after herbicide treatment ofCarex oxyandra communities. At the early stage ofestablishment, neither germination of seeds nor
survival for half a year past germination weresatisfactory. Half a year after sowing 400 seeds per
m2, at most 22.2 individuals per m2 (5.6%) could becounted in fall 2010 (Table 1). This percentage isinferior to the commonly expected germination
percentage (from 10 to 40%) of Japanese cypress
seeds [9].
One of the causes of inferior establishment is
considered to be the thick mat of withered sedge
leaves, which occupied the soil surface for about 3
years in the present study. Removing such a thickobstacle that interferes with root elongation into thesoil is critical. Japanese cypress seeds are viable fora short period [9], so they cannot survive until themat decomposes. It is probable that more seeds had
germinated initially, but most died and had
disappeared before our survey, which also shows theseverity of the growth environment. A gradualdecrease in the number of individuals and elongationin height after 2011 indicates a phenomenon like
self-thinning, which could be caused by an inferior
environment in the area of the withered sedge, rather
than by competition among Japanese cypress
seedlings. Therefore, withering of a Carex oxyandra
community is insufficient to promote theestablishment of seedlings.
Considering the inferior early growth of the tree
seedlings, some additional treatment is needed to
improve the growth environment. The mostcharacteristic Carex oxyandra community after
herbicide treatment is a thick, carpet-like mat ofwithered leaves. It has been reported that soil
disturbance plays an important role in increasing thesurvival rate of viable seeds of Japanese cypress [10],and soil scarification to remove dwarf bamboo iseffective in promoting regeneration of a forest(though the effect depends on the physiochemical
features of the soil) [13].

REFERENCES[1] Akai, T. and Asada, S., “Studies on Natural
Regeneration (I) The reproduction of
Chamaecyparis at wet podozolic zone in Kiso
District”, Bulletin of the Kyoto University
Forests, Vol. 39, 1967, pp. 35-63. (in Japanesewith English Résumé)[2] Akai, T., “Studies on Natural Regeneration (II)the regeneration of Chamaecyparis obtusa at
Miure Experimental Forest in Kiso District”,
Bulletin of the Kyoto University Forests, Vol.
44, 1972, pp. 68-87. (in Japanese with English
Résumé)[3] Suzaki, T., “Study on herbicidal use insilviculture”, Bulletin of the Kyushu University
Forest, 42, 1968, pp. 99-233 (in Japanese withEnglish Résumé)[4] Hasegawa,
M.,“Effectofdifferent
regeneration methods on species compositionand sapling density of regenerated forests
within clear-cut areas in a heavy snow region”,
Japanese Journal of Forest Environment, Vol.
47 (1), 2005, pp. 9-20 (in Japanese with English
summary)[5] Aiura, H., “Erosive phenomenon in heavy snow
mountainous regions in Japan”, Journal ofForest Research, Vol. 7 (2), 2002, pp. 57-66 (inJapanese with English summary)[6] Takamatsu, T., Kohno, T., Ishida, K., Sase, H.,
Yoshida, T. and Morishita, T., “Role of thedwarf bamboo (Sasa) community in retaining
basic cations in soil and preventing soil

CONCLUSIONSIn the present study, the effect of herbicide was
examined at various dilution levels on suppressing
the community of Carex oxyandra in the Miure
Experimental Forest within the Kiso National Forest.
Coverage and plant height of Carex oxyandra
decreased obviously after a half year in all dilution
530GEOMATE- Bangkok, Nov. 14-16, 2016

floor”, Journal of the Japanese Forestry Society,vol. 66 (12), 1984, pp. 483-490 (in Japanesewith English summary)[11] Wootton, L.S., Halsey, S.D., Bevaart, K.,
McGough, A., Ondreicka, J. and Patel, P.,“When invasive species have benefits as well as
costs: managing Carex kobomugi (Asiatic sandsedge) in New Jersey’s coastal dunes”,
Biological Invasions, Vol. 7, 2005, pp. 10171027[12] Arase, T. and Uchida, T., “Growth andpersistency of five terrestrial sedge species on a
cutting slope”, Journal of the Japanese Societyof Revegetation Technology, Vol. 35 (1), 2009,
pp. 119-122 (in Japanese)[13] Yamada, K., “Growth of secondary forests onserpentine soil after different soil disturbanceintensity”, Journal of the Japanese Forestry
Society, Vol. 81 (4), 1999, pp. 291-297 (inJapanese with English summary)acidification in mountainous areas of Japan”,
Plant and Soil, Vol. 192, 1997, pp. 167-179[7] Morisawa, T., Sugita, H., Hashimoto, R. and
Akai,
T.,
“NaturalregenerationofChamaecyparis obtusa for 36 years by Strip
logging in relation to dwarf bamboo elimination，
examined by aerial photographs in Kiso”,
Journal of the Japanese Forestry Society, Vol.
92, 2010, pp. 22-28 (in Japanese with English
summary)[8] Arase, T., Okano, T. and
Shirota, T.,“Colonization and morphological changes of asedge restricting the regeneration after winddamage In a natural forest in Kiso District,central Japan”, Sixth International Conference
on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 2016
(under submission)[9] Nakayama, K., Seed germination of usefulplants. Tokyo: Uchida Rokakuho Publishing Co.
Ltd., 1966, pp.282-285 (in Japanese)[10] Yamamoto, S. and Tsutsumi, T., “Thepopulation dynamics of naturally regenerated
hinoki seedlings in artificial hinoki stands (III)Dynamics of the seed population on the forest531

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DYNAMICS OF EXOTIC GRASS COVER PLANTS ON SLOPES
Taizo Uchida1, Jun Tanaka2, Kentaro Kondo3, Daisuke Hayasaka4, Yuki Tomoguchi3, Teruo Arase5 andTetsuo Okano5
1
Faculty of Engineering, Kyushu Sangyo University, Japan; 2Conservation Engineers Co., Ltd.,
3
Japan; Graduate School of Engineering, Kyushu Sangyo University, Japan; 4Faculty of Agriculture, Kindai
University, Japan; 5Faculty of Agriculture, Shinshu University, Japan

ABSTRACT
Alien species in the family Poaceae play an important role as cover plants for erosion control on slopes,
particularly in Japan. However, concerns have arisen regarding the adverse effects of these species on the local
ecosystem and biodiversity. This study examined the succession and seed propagation of alien Poaceae that areused for erosion control on the cut slopes of Sakurajima volcano in southeastern Japan. Although the alien
species used as cover plants were dominant for 2-3 years following their introduction to the slopes, they weredisplaced entirely by native species after 6 years, which is considerably faster than ordinary succession on othervolcanoes. In addition, heading (flowering) of alien species were rarely observed. It is considered that theobserved results were caused by the oligotrophic conditions that resulted from the bioengineering techniques thatwere employed on Sakurajima.
Keywords: Alien, Cover plant, Cut slope, Erosion control, Poaceae, Seed propagation, Succession, Volcano

after being introduced, all of the alien species in thefamily Poaceae that were used for erosion control on
Sakurajima volcano were completely displaced bynative plant species. Although they suggested thatthe potential of these alien plants to disperse had
decreased in this area, the risk of seed dispersal bythese species in this six-year period has not yet beenassessed. We therefore reinvestigated the successionof alien plant species in the family Poaceae that areused as cover plants for erosion control on
Sakurajima and examined the seed-propagation
ability of these species.

INTRODUCTION
There are numerous active volcanoes around theworld. The slopes of many of these volcanoes are
typically bare or sparsely vegetated due to volcanic
ejecta, such as pumice, scoria, ash, and volcanic gas,causing soil erosion and debris/mud flows.
Consequently, volcano slopes are intentionallycovered with plants for erosion control. In Japan,
numerous alien species in the family Poaceae havebeen used for this purpose, including Agrostis
stolonifera, Cynodon dactylon, Dactylis glomerata,
Eragrostis curvula, Festuca arundinacea, Festuca
rubra, Lolium multiflorum, Lolium perenne,
Paspalum notatum, Phleum pretense and Poa
pratensis. These species, which are indigenous toEurope, South Africa and South America, are well
suited to growing in Japan where they typicallygrow at much faster rates than the native species.
However, concerns have arisen regarding thenegative effects that these alien plant species haveon the ecosystems and biological diversity present in
the areas that they invade. Although numerous
studies have been conducted on plant succession on
volcanoes, e.g. [3], [5], [6], [8], [13], [15]-[17], [24][31], [33], [36], [37], only a few have examined thesuccession of alien species used as cover plants for
erosion control on volcanoes [14], [18], [20], [21],[34]. It is therefore important to further clarify the
succession of alien plant species on volcanoes forpreventing soil erosion and debris/mud flows, andconserving ecosystem integrity and biologicaldiversity.
Kondo et al. [14] reported that within six years

METHODS
Study Site
The archipelago of Japan, which is located on the
Pacific Ring of Fire, has as many as 110 active
volcanoes [11]. Of these, Sakurajima in Kagoshima
Prefecture in southwestern Japan, which is where
this study was performed, is one of the most active.
Sakurajima is located in the warm temperature zone
where the mean annual temperature and annual
precipitation over the last decade (2006-2015) haveranged from 18.2 to 19.3°C and 1,530 to 3,664 mm,
respectively. Volcanic activity accompanied by ash
fall, soil erosion and debris/mud flows continue tothis day.
In this study, the 85 cut slopes that resulted fromthe construction of check dams were surveyed on theeastern side of Sakurajima; these sites included someof the same sites investigated by Kondo et al. [14].
The slopes, which were created during 2005-2014,
532

GEOMATE- Bangkok, Nov. 14-16, 2016

were located at an altitude of 366 to 655 m above
sea level and their angles of inclination ranged from25 to 82°. Volcanic gas appeared to have no effecton the vegetation on the cut slopes, but ash-fall wasconstantly observed on the slopes.
Alien species belonging to Poaceae, such as A.
stolonifera, C. dactylon, D. glomerata, E. curvula, F.
arundinacea and F. rubra, have been introduced for
erosion control on the cut slopes in this area using abioengineering technique known as the slurry
application method, which is an aerial seeding
method. While the method typically calls for the
repeated application of additional fertilizer [12], [18],[22], no additional fertilizer has been applied to theslopes of Sakurajima.

aaa

n=85
Cynodon dactylon

α (n=10)

Dactylis glomerata
Festuca rubra

β (n=40)

Dactylis glomerata

Festuca rubra

γ (n=7)

Miscanthus sinensis

δ (n=28)

Miscanthus sinensis
Polygonum cuspidatum

Vegetation Surveys

Rhaphiolepis indica

Surveys of the vegetation on the cut slopes wereconducted in October 2015. Three survey quadrats(1 × 1 or 2 × 2 m) were randomly placed on each cut
slope. The species composition of the plants in eachquadrat was recorded and scored using the BraunBlanquet cover-abundance scale [2]. All of therecorded plant species were categorized as native or
alien, and as ferns, herbaceous plants or woody
plants based on published literature [1], [10], [19],[23]. In addition, heading (flowering) of Miscanthus
sinensis (native) and F. rubra (alien) was also
observed at this time, and the number of scapes ineach quadrat was also enumerated.
The number of years that had passed since thebioengineering techniques were implemented(NYPSB) was obtained from the Kagoshima District
Forest Office, Kyushu Regional Forest Office,Forestry Agency. The Braun-Blanquet coverabundance scale (r, +, I, II, III, IV, and V) wastransformed as follows: r and + were taken as 0.1%;
I as 5.0%; II as 17.5%; III as 37.5%; IV as 62.5%,
and V as 87.5%. The cut slopes were classified into
vegetation types based on their species composition
by a two-way indicator species analysis,
TWINSPAN, using the PC-ORD statistical software
package (ver. 4.0 for Windows, MjM Software
Design, OR).

1

2
Division level

Fig. 1

Fig. 1

3
Types

Classification of vegetation on cut
slopes by TWINSPAN, using species
composition (coverage data, %).
Cut levels of 0, 20, 40, 60, and 80
were employed in the analysis.

used as cover plants for erosion control and reached82.17% in coverage (Table 1). Type β plants were
dominated by the alien species F. rubra and D.
glomerata, which are also used for erosion control.
However, the extent of coverage by these alien
species was lower (62.18%), while that of native
species was higher than it was for Type α plants. For
Type γ plants, coverage of alien and native species
was generally similar; for example, coverage for F.
rubra and M. sinensis was 10.26% and 17.38%,
which was the highest coverage for the alien andnative species of this type, respectively. Conversely,
Type δ vegetation was characterized as having
almost no alien species (coverage <0.01%;
frequency was 2/28 slopes) and native species,
mainly M. sinensis (53.51%), were dominant.
The NYPSB for each type is shown in Table 1.
In terms of succession, the findings show that thevegetation communities on cut slopes would likely
progress through types α, β, γ and finally δ after the
bioengineering technique had been implemented.

RESULTS
Vegetation TypeThe vegetation on the 85 cut slopes wasclassified into four types (α, β, γ and δ) using
TWINSPAN (Fig. 1); types α, β, γ and δ were foundon 10, 40, 7 and 28 slopes, respectively. M. sinensis,
D. glomerata, F. rubra, Polygonum cuspidatum,
Rhaphiolepis indica and C. dactylon were used asindicator species.
Type α plants were dominated by the alien
species C. dactylon and F. rubra, which had been

Relationship between NYPSB and Component
Species
The relationship between NYPSB andcomponent species is shown in Fig. 2. At two years
after the bioengineering technique had been
implemented, most of the species on the cut slopes
533

GEOMATE- Bangkok, Nov. 14-16, 2016

aaa

Table 1 Plant coverage for each vegetation type classified
aaaaa by TWINSPAN.
Vegetation types
α(n=10)

Speciesβ(n=40)

γ(n=7)

δ(n=28)<Alien species>
Cynodon dactylon #, †

63.17

2.92

0.96

0.01>

Festuca rubra #, †

19.00

30.09

10.26

0.01>

Dactylis glomerata #, †

3.77

32.09

Trifolium repens #

0.51

2.07

0.24

0.01>

Phalacroloma annuum #<Native species>
Miscanthus sinensis #Polygonum cuspidatum #0.01>

Alnus firma ω

1.54

17.38

53.51

0.01

0.02

2.65

0.01>

0.01

1.23

Eurya japonica ω
0.05

Artemisia indica #0.01>

Lespedeza cuneata #

0.01>

1.33

0.01>

0.06

0.01>

Pinus thunbergii ω

0.39

Rhododendron obtusum ω

0.06

Rhaphiolepis indica ω

1.05

Albizia julibrissin ω

0.01>

Boehmeria spicata #0.01>

Dennstaedtia hirsuta θ

0.01>

Indigofera pseudotinctoria ω

0.01>

Ligustrum japonicum ω

0.01>

Frequency ‡

10/10

40/40

7/7

2/28

NYPSB (Min.-Max.)

1.8
(1-2)

2.0
(1-4)

3.4
(3-5)

6.6
(4-10)#, herbaceous plant (excluding fern); †, species (Poaceae) used as cover plants for erosion control; ω, woody plant; θ, fern and‡, number of slopes which had alien species/total number of slopes. See Fig. 1.

were alien species. Of these, almost all (98.4%) were
species that had been used as cover plants for
erosion control (i.e., C. dactylon, F. rubra and D.
glomerata). However, the coverage of these species
decreased markedly after the third year, before
disappearing completely by the sixth year after
implementation of the bioengineering technique.
Native species were observed from the first year,
and M. sinensis and woody plants were observed
from the third year onwards; these species became
dominant after the fourth year, particularly M.
sinensis (83.7%).

technique, but they were not observed by the fourth
year; specifically, the number of F. rubra scapes
was 0.1 m-2 in the first year, 1.8 m-2 in the second
year, and 0.2 m-2 in the third year, which aresignificantly low compared to those for M. sinensis
at the experimental sites, and general F. rubra.
DISCUSSION
It is considered that implementation of thebioengineering technique would result in thecomposition of the plant communities on the cut
slopes of Sakurajima progressing gradually from
communities dominated by alien Poaceae, whichhave been used extensively as cover plants for
erosion control, to communities dominated by native
species (Fig. 1 and Table 1). In this study, native
species became dominant after the fourth year andalien species disappeared entirely by the sixth year
(Fig. 2). As also reported by Kondo et al. [14], therate of alien plant succession on Sakurajima was

Seed Propagation
Scapes of M. sinensis were observed three years
after implementing the bioengineering technique(Fig. 3). Scape numbers increased gradually
thereafter, reaching 22.0 m-2 by the tenth year.
Festuca rubra scapes were observed in the firstyear after implementation of the bioengineering
534

GEOMATE- Bangkok, Nov. 14-16, 2016

markedly higher than succession on other active
volcanoes where alien species have remained
established for more than 10 years (10-33 years)
after implementation of the bioengineering
technique [18], [20], [34]. In addition, scapes ofalien species such as F. rubra, which have been used
extensively as cover plants, were rarely observed on
Sakurajima (Fig. 3).
It is therefore considered that alien species used
for erosion control on Sakurajima have little effecton ecosystem integrity and biological diversity whencompared to other active volcanoes [18], [20], [34];however, we cannot state with certainty that thesealien species will not disperse to other areas, andfurther investigation is needed. Meanwhile, althoughalien species disappeared rapidly on the slopes of
Sakurajima, their coverage was initially relatively
high after implementation of the bioengineering
technique (Fig. 2). It is therefore considered thatthese plants played an important role in erosion
control on Sakurajima. Indeed, none of the slopes
examined in this study showed any evidence oferosion.
By the way, why did all of the alien species in
the family Poaceae rapidly disappear on the cut
slopes of Sakurajima? Since the nutrient (fertilizer)
demand of alien species in the family Poaceae istypically very high [4], [7], [9], [32], [35], the alienaaaplants on Sakurajima would initially have flourished
due to the eutrophic conditions that would haveaccompaniedtheimplementationofthebioengineering
technique.
Conversely,
theaaaaaaaaaa

Number of scapes (No. m-2 )

30.0

20.0

Miscanthus sinensis

10.0

Festuca rubra

0.0
2

4

6

8

NYPSB

Fig. 3

Changes in the number of scapes after
implementation of the bioengineering
technique.
Error bars indicate the standard deviation.

Fig. 3Cumulative coverage of species (%)

100

75

50

25

0
1

2

3

4

5

6

7

8

9

10

NYPSB

: Alien (Cynodon dactylon) †: Native (Miscanthus sinensis): Alien (Festuca rubra ) †: Native (woody plant): Alien (Dactylis glomerata) †: Native (others): Alien (others)

Fig. 2

10

Relationship between NYPSB and component species on the cut slopes of Sakurajima.
†, species (Poaceae) used as cover plants for erosion control.
535

GEOMATE- Bangkok, Nov. 14-16, 2016

subsequent decrease observed in alien plant
coverage may have occurred rapidly due to theestablishment of oligotrophic conditions resulting
from no more fertilizer being applied to the slopes.
These findings would be supported by reports ofalien species at other active volcanoes being present
more than ten years after being introduced, where
fertilization schemes including the use of a thickcultivation base, additional fertilizer at a later date
and arbuscular mycorrhizal fungi, were employed[18], [20], [34]. Incidentally, of C. dactylon, F.
rubra and D. glomerata which have been used ascover plants for erosion control, C. dactylon iswarm-season grass, and the rest belong to coolseason grass. It seems therefore that the difference
between the two is not directly related with theabove succession of alien species on the cut slopesof Sakurajima (see Table 1 and Fig. 2).
It is not clear why heading (flowering) of F.
rubra were rarely observed on the slopes of
Sakurajima, but we consider that volcanic ejecta
such as ash might have some influence seedpropagation of alien species in the family Poaceae ina negative way.
The high germination rate and rapid growth ofalien plants in Japan mean that using alien species ascover plants is indispensable for erosion-control
efforts in Japan. In addition, sales of these species
are already well established and they can beobtained more easily and cost more effectively thanindigenous Japanese species. However, as theconcerns have increased in recent years regarding
the adverse impacts of alien species on ecosystems
and biological diversity, accumulating knowledge on
the invasiveness of alien species as cover plants isimportant for developing future erosion control
measures. Thus we consider that findings of thisstudy on the succession of alien plants used for
erosion control on the slopes of Sakurajima would
be very meaningful.

their succession on the slopes of Sakurajima. It wasalso clarified that heading (flowering) of alien
species, F. rubra, which have been used as coverplants for erosion control were hardly formed after
implementing the bioengineer technique on the cut
slopes of Sakurajima.
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CONCLUSIONS
Our study provides information on the
succession of alien Poaceae, which have been used
as cover plants for erosion control on the cut slopesof Sakurajima volcano. Although alien plants used
for erosion control were dominant for the first
couple of years from implementation of thebioengineering technique, native species were
dominant thereafter and alien species completely
vanished by the sixth year, which corresponded wellwith previous report by Kondo et al. [14]; this trendof succession on the slopes of Sakurajima would
have been caused by fertilizer management (i.e. low
fertilizer application) employed on Sakurajima. In
contrast, the difference between grass types (warmor cool-season grass) of alien species used for
erosion control would not likely have an effect on
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Sakagami N, Collado M & Oguchi T, 2011.
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Pinatubo, the Philippines. Journal of Geography
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M, Niimi Y, Ozaki A, Ohkubo T, Ueki T,
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Unzen-Fugen. J. Jpn. Soc. Reveget. Tech. 32,
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Komagatake, Hokkaido. 1. Akaikawa pumiceflow. Ecological Review 16, 271-292.

[14] Kondo K, Uchida T, Hayasaka D, Tanaka J,
Sato A & Arase T, 2016. Vegetation succession
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erosion control, Mt. Sakurajima. International
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Moscow
University
Biological Sciences Bulletin 70, 99-103.
[16] Korznikov KA, 2015. Vegetation cover at the
Maguntan mud volcano (Sakhalin Island ，
Russia): species composition and spatial
distribution. Phytocoenologia 45, 125-134.
[17] Marler TE & del Moral R, 2013. Primarysuccession in Mount Pinatubo. Habitat availability and ordination analysis. Communicative& Integrative Biology 6, e25924 (online).
[18] Miyasihita S & Yamada M, 2010. Succession ofvegetation on the southern slope of Mount Usu
after aerial seeding work. Forest Consultant 122,
8-14. In Japanese[19] Miyawaki A, Okuda S & Fujiwra R, 1994.
Handbook of Japanese vegetation. Shibundo,
Tokyo. 646 pp. In Japanese[20] Nakano Y & Futami T, 2004. Observation onplant succession at the past cut slope
landscaping sites in Miyakejima Island. J. Jpn.
Soc. Reveget. Tech. 30, 383-388. In Japanese[21] Ogawa Y, Akema T & Daimaru H, 2011. A
field survey of revegetation plants and fieldobservation of overland flow on slopes overlain
by pyroclastic-flow deposits, Unzen volcano,Japan. Journal of the Japan Society of Erosion
Control Engineering 63, 78-82. In Japanese[22] Ogawa Y, Shimizu A, Shimizu T, Daimaru H &Miyabuchi I, 2002. Surface runoff and sediment
discharge for three years on a slope revegetatedby aerial seeding work at Unzen Volcano. J. Jpn.
Soc. Reveget. Tech. 28, 255-258. In Japanese[23] Shimizu T, 2003. Naturalized plants of Japan.
Heibonsha, Tokyo. 337 pp. In Japanese[24] Tagawa H, 1964. A study of the volcanic
vegetation in Sakurajima, south-west Japan. I.
Dynamics of vegetation. Mem. Fac. Sci.
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[25] Tsuyuzaki S, 1987. Origin of plants recoveringon the volcano Usu, northern Japan, since theeruptions of 1977 and 1978. Vegetatio 73, 5358.
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537

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

NEW DATA ON AFRICAN-EURASIAN LIMIT AND SPATIAL
DISTRIBUTION OF THE CURRENT DEFORMATION THROUGH
ACTIVE STRUCTURES IN THE CENTRAL MEDITERRANEAN AND
THE SURROUNDING AREAS
Adnène KASSEBI 1,2 and Fouad ZARGOUNI 1
1

Faculty of Sciences of Tunis, University of Tunis El Manar, Tunisia; 2 National Center for Mapping andRemote Sensing, Tunis, Tunisia

ABSTRACTThe Mediterranean region is constituted by a complex mosaic of plates and microplates, which move relative
to each other. Although there is general agreement to consider that the subduction and closure of the Tethyan
Ocean and the interaction between European and African dishes were the main drive mechanism controlling theevolution of basins orogenic extension and arches in the western and central Mediterranean.
We aim in this work, introduce new insights into the evolution of time and space scattered relics of theexternal areas of the maghrébide belt in the central Mediterranean, and its tectonic behavior as part of the current
border of the Africa-Europe course.
The integration of digital data (in a Geological Spatial Database) to seismic processing and spatio-temporal
data analyzes performed, allows the identification of new data on seismic risks and Hazards associated withzones of tectonic surface and deep. Trading volumes in our project are directly related to geometric, topographic
and geological characteristics of the study area and the characteristics of lineaments that integrates data mining
methods and different geoprocessing made.
We define segment boundaries internal offshore zones, and characteristics of tectonic structures responsible
for their arrangement, and we finally found those who are still working. The flush results are: (i) the kinematic
active deformations in the Mediterranean area; (Ii) the spatial distribution of the current deformation throughactive structures in the central Mediterranean (Sicilian-Tunisian Strait and surrounding lands); (Iii) the sliding
velocities of seismic recurrence and associated risks.
Keywords: Central Mediterranean, Seismic processing, Active fault, Spatio-temporal data, Earthquakes
Western Mediterranean.
The distribution of focal mechanisms ofearthquakes such as the various types of deformities
found in the Mediterranean area suggest that the
distribution of deformation zones (that is to say,affected by a single type of deformation zones) does
not follow a simple rule and that only theinterpretation of geological and geophysical data
may lead to a coherent seismotectonic a complex
domain model. To define the distribution and spatial
variation of various types of deformation as well as
the evolution of the stress field from the Pliocene.
The example we have considered is the Tyrrhenian
Arc and Tellian extension in northern Tunisia which
seems to correspond to a transition from subduction
to crash diets.
2. GEOGRAPHICAL SETTING OF THE
STUDY AREA

1. INTRODUCTION
Earthquakes in the Mediterranean, recorded inrecent years strong Ambrasseys, 1962; Rothe, 1969;
McKenzie, 1972, 1978; Udias et al., 1976; Cagnetti
et al., 1976; Hatzfeld, 1978; Bonjer and Fuchs,
1979; Mayer-Rosa and Pavoni, 1977; Ben Ayed,
1986; Zargouni and Abbès, 1987; Zenati, 1990;
Boutib and Zargouni, 1997; Dlala, 2002 ; Kacem,
2004; Kassebi, 2007, Kassebi et al., 2013, Kassebi
and Zargouni, 2013, 2014 and many neotectonic
studies in the Mediterranean area clearly show thatvery different deformation mechanisms can coexist
in areas very close to one than another. But thedifferent types of distortion to be the expression of
the same phenomenon; in this case, the N-S
convergence of European and African plates acting
for the past 70 Ma (Pitman and Talwani, 1972,
Dewey et al, 1973. Le Pichon et al, 1977.
Tapponnier, 1977 ; Savostin et al. 1986). This
convergence, which is stronger in the eastern part of
the Mediterranean (McKenzie, 1972) is absorbed by
the deformations that are distributed in the regionsof variable size: near the Aegean and Tyrrhenian
subduction zone and continental collision zones

The central Mediterranean and specifically the
Sicilian-Tunisian Strait composed by shoals locatedbetween Sicily and Tunisia, also called SicilianTunisian threshold is a sea separating the two shoals
or basins of the Mediterranean Sea: 1st, consisting of
the Algerian-Provencal basin and the Tyrrhenian sea
538

GEOMATE-Bangkok, Nov. 14-16, 2016

forms the western Mediterranean; the 2nd consistingof the Ionian Sea and the Levantine basin forms theeastern Mediterranean. The first covers an area of
about 0.85 million square kilometers while thesecond covers about 1.65 million square kilometers.
Both sets shallows form abyssal plains with a depthof 2000 to 5000 m and consist of an ocean floor.
This stretch of sea is bordered to the northwest
by Sardinia, to the south by the Pelagie Islands(including the island of Lampedusa) and southeast of
the islands of Malta which, together with Sicily,
Malta channel ; the latter channel is sometimes
geopolitically integrated as part of the broader Sicily
Channel to designate the whole area south of Sicily
in Italian territorial waters.
In its narrowest part, between Cap Feto near
Mazara del Vallo and near Cap Bon-El Haouaria, thecanal is 145 km wide.

other (Fig. 1).
North Africa is bound by an Alpine orogeny kind
resulting from the subduction and closure of the
Tethyan Ocean and the interaction between theEuropean and African dishes. Sicily lies on part ofthe complex convergent boundary where the African
Plate is subducting beneath the Eurasian Plate. Thissubduction zone is responsible for the formation of
the stratovolcano Mount Etna and considerable
seismic activity. Most damaging earthquakes
however, occur on the Siculo-Calabrian rift zone.
This zone of extensional faulting runs for about 370
kilometres (230 mi), forming three main segments
through Calabria, along the east coast of Sicily and
immediately offshore, and finally forming thesoutheastern margin of the Hyblean Plateau. Faults
in the Calabrian segment were responsible for the1783 Calabrian earthquakes sequence (Catalano et al.
2008).
In Tunisia, as in the rest of the Maghreb, recent
strains are related to the collision (Bousquet, 1977;
Ben Ayed and Viguier, 1981). The examination ofextensional tectonic compressive and the Alps
Tunisia and its foreland shows that these tectonic are
not any. In particular, the compression anddistension present a orthogonal directions extension
and shortening. Indeed, the arrangement of
deformations and quaternary neogene Tunisia,
shows that in general, the maximum principal stresswhich is either σ1, σ2 either, keeps almost the same
direction.
Tectonics first located only in the Far North of
Tunisia and is associated with the episode ofestablishment of groundwater is contemporaneous
strike-slip extensional tectonics central Tunisia
ahead of the front of tablecloths. Over time, thecompression zone has migrated to the South; it willreach the central Tunisia during the lower Tortonian.

Fig. 1 Geographical and geological map of the studyarea.
3. GEOLOGICAL HISTORY
The central Mediterranean Sea separated from
the Messinian evaporites the eastern and the western
Mediterranean basins (Fig. 1). The Sicilian-Tunisianthreshold is of major interest for the study of sea
level changes, exchange of seawater, sediment andtransits the importance of the carbonate production
during the pre-evaporite history of these areas. This
threshold zone is formed by the Pelagian platform,
extending the Mesozoic and Cenozoic of southern
Sicily, Tunisia and Tripolitania, extending
northward Saharan craton. It is now almost
completely submerged in shallow (Winnock & Bea
1979 Pedley & Grasso 1992). This bedrock was
deeply sawed Pliocene by a network of faults N120
in response to crustal thinning axis maximum in thearea of Pantelleria (Reuther 1987) WNW-ESE,
equated to an area of bulge pre-rift age upper
Miocene, largely collapsed in the Plio-Quaternary.
Thus, the few witnesses to the Messinian
sedimentation they are only limited to areas ofoutcrop horst now emerged, that of the Maltese
Islands on the one hand, on the island of Lampedusa

4.
GEOMETRICANALYSISANDSTRUCTURAL

4.1. Structural analysis
The presence of three key unconformities isimportant to note and is the key to unravelling thepresence and location of traps. (Messinian, IntraTortonian and Pyrenean)•
The
Pyrenean(Oligocene)
Unconformity is present and marks a change instructural style in the deeper section. In well w1, thisis found just above the Bou Dabbous Formation.
• Understanding the biostratigraphy is at times
more important that the lithostratigraphy as
names change across the border and from well to539

GEOMATE-Bangkok, Nov. 14-16, 2016

well.
• The “w2” sandstones and the “w3” Carbonate(Fig. 2) are most likely time equivalent units
encapsulated in the Mahmoud Shale. Just
different depositional environments.
• The Nummulites Fichteli (marks the top of Bou
Dabous) whereas the Nummulites Vascus (marks
the top of the Soaur / Base of Fortuna and end of
the Pyrenan phase).
An alternative interpretation: “w2” and “w3” arenot Formations (Fig. 2). These events are
stratigraphic units (i.e. w2 sands and w3 limestones)
representing different depositional settings during
the Mahmoud Formation. The key is thebiostratigraphy.
It is then overprinted at times by long N-Slineaments representing the extensional phase.

Fig. 5 Regional cross section NO. 3
1. Two main structural trends are observed in
Sicilian-Tunisian threshold as extended from theSouth and North of the permit:- NW-SE Normal Faulting (possibly MessinianTime);- SW-NE Thrust Faulting (possibly Pre-Pyrenean
Time);
2. There is a third overprint of the NNW-SSE
reidel shears as seen on the 3D;
3. Several trap styles (both shallow and deep)
could be assumed;
This structural style above in the shallow anddeep can be seen in the shallow and deep throughout
the Sicilian-Tunisian threshold.
It is then overprinted at times by long N-Slineaments representing the extensional phase.
4.2. Geodynamic analysis

Fig. 2 Regional cross sections

A special comment must be devoted to theevolutionary model of the western Mediterranean
region given by Dewey et al. (1989), since it is thelast proposed in literature and in addition, it presents
the most detailed reconstruction of the AfricanEurasian kinematics during the time that we
considered.
This model suggests, among other things, thatthe migration of the Calabrian Arc (to the Southeast),
and the consequent opening of the Tyrrhenian basin,
occurred in the context of African-Eurasianconvergence led just against it (SE-NW). No clear
explanation is given dynamics of this very special
phenomenon. Only a vague reference is mademechanisms back-arc or rolling back.
The final note of Dewey and colleagues about
the evolutionary models in this area is a direct
relationship cannot always be provided between therelative movement of blocks of imprisonment andmain trends deformations inter dish. They attribute
this mismatch to the complex kinematics of microplate in the central Mediterranean area. We believe,
instead, that this problem is due to the fact that the
post-Tortonian kinematics of Africa-Eurasia
assumed by Dewey et al. (1989) is not reliable. Thishypothesis is also suggested by a strange detour into

Fig. 3 Regional cross section NO. 1 Casero, 2004

Fig. 4 Regional cross section NO. 2

540

GEOMATE-Bangkok, Nov. 14-16, 2016

the path of movement of Africa proposed by Deweyet al., the drastic change of the drifting trend of thisdish, from NE to NW. Such rapid movement of large
deflection of continental blocks does not seem very
likely the physical point of view and, in addition, it
is supported by any clear change in pattern of
deformation in the areas around the AfricanEurasian system and in the Mediterranean region inparticular.

these are a potential hazard for the coastal cities ofTunisia and Sicily. Only the SW–NE striking
structures seem active in the present-day stress field,
whereas NW–SE structures are apparently not.
These latter ones were probably active under adifferent stress field. Some authors suggest thepresence of active strike-slip faults along theSicilian-Tunisian strait.
However, according to the recent focal
mechanisms and seismicity distribution, it appears
that these faults have no clear activity today. They
may have played an important role in the past (atleast before the Quaternary, but probably before thebeginning of the compressive stage of the margin,
during the Miocene), and some of them favourablyoriented could be reactivated as thrust faults in thepresent-day stress field. This could be the case of theborder faults of the horst Pantelleria (Fig. 4).
Conversely, the series of north-verging thrusts
offshore Tunis Gulf is probably neoformed PlioQuaternary structures, and corresponds to thebeginning of the margin inversion. The Cap Bon
area is slightly seismogenic, with small earthquakesthat appear spread offshore, according to the INM
catalogue of earthquakes from 1971 to present. Most
large earthquakes occurred inland, and areassociated to Tellian folds and thrusts, some otherexamples of more limited active features are foundoffshore.

Fig. 6 Time Structure Cretaceous Abiod Contour
Interval 100msec.
4.3. Discussion
4.3.1. Limits of the Internal Zones
From the observations described in the sections
above, we propose a new structural map for theoffshore domain (Fig. 6). With the deep coring,basement samplings and dense seismic reflection
data in order to support our hypotheses, we attempt
to relate the lithology offshore to the one onshore
from the changes of geometry, morphology,
bathymetry, roughness and seismic structures
reported.
The contact between Internal and External Zones
is clearly mapped at east of Bizerte, along the east–west accident delimitating the southern end of the
Kechabta bassin deep-sea fan (Fig. 6). Elsewhere,
we have used several geomorphic markers andchanges in the roughness of the bathymetry orseismic facies, together with geological evidence
onland, in order to propose a connection of offshore
domains either to the Internal Zone or External
Zones. However, several limitations and doubts
about the basement nature remain, that are discussed
hereafter.

5. SEISMIC DEFORMATION
Literature review of work on the seismicity of
the central Mediterranean seismic analysis shows
more or less variable and sometimes contradictory,
the authors suggest a quiet seismicity in Tunisia as
reliable sources shows destructive earthquakes ofregistration in Part north-eastern, central and in
offshore Tunisia (Fig. 7).

Fig. 7 Seismicity map of Sicilian-Tunisian Strait

4.3.2. Reactivation of structures
The seismicity from the last three decades shows
great changes from east (Pantelleria Rift zone) towest (Tunis gulf zone). The Pantelleria Rift area isby far the most active. Nowadays, the major active
faults depict compressional focal mechanisms, andComparing total and seismic deformation along
the Northern Africa to Sicily sections, we find aninteresting relation between geometry and amount of
deformation due to seismicity. Within sections
affected by pure thrust motion we find a higher
541

GEOMATE-Bangkok, Nov. 14-16, 2016

seismic deformation (26% for the Algeria section
and 10% for the Tunisia to Sicily section) withrespect to the sections where the strike-slip motion
prevails, characterized instead by very small
percentages of seismic deformation (<1%) (Pondrelli,1999).
For all sections, however, the seismicity accounts
for low percentages of total expected deformation
normal to the boundary: 31% in the Calabria section,
30% in the Central to Southern Apennine section,
3% in the Friuli section and 10% in the Eastern
Adriatic section. Only in the Sicily Strait section do
we find a high value of strain accommodated byseismicity that here accounts for 79% of total strike
slip deformation (Pondrelli, 1999).
It is clear that these values are generally low, but
comparable values have also been found in this area
(Jackson and McKenzie, 1988; Pondrelli et al.,
1995; Ward, 1998). We assume that these small
percentages are due to the fact that 100 years are notrepresentative of the seismic recurrence cycle ofthese zones, as already suggested by Ward (1998).
However, more and more studies are showing thataseismic deformation is effectively high and often
related to silent events (or slow earthquakes)(Amoruso et al., 1998; Hirose et al., 1998; Linde etal., 1996, 1998).

where the slab is removed and the opening of the
Tyrrhenian basin product migration to the East
(Malinverno and Ryan, 1986) and the likely rotationof the plate Adria (Anderson and Jackson, 1987).
The consequent NE-SW prevailing strain patternoverprints the NNW-SSE Africa Eurasia plateconvergence (Pondrelli, 1999).
7. CONCLUSIONSThe summation moment tensor has been appliedin the western Mediterranean from data representing
100 years of intermediate to strong earthquakes (3<M). Percentages ranging from 3 to 31% of strain
received by seismic activity have been found. Only
the section of the Sicilian-Tunisian Strait made thestrike slip seismic deformation account for 79%(Pondrelli 1999).
We still do not know if the seismic deformation
can reach such high values, and we must rememberthat 100 years may not be representative of theseismicity recurrence characteristic of the central
Mediterranean. In all cases, the amount of seismic
deformation is lower than expected, but the
deformation pattern obtained has some interesting
features that the movement of the overall plate andgeodetic measurements do not disclose, astranspressional model along Africa North to Sicily
area or the influence of the evolution of the periTyrrhenian area on this deformation periadriatic
sections.
The northwest side of the Sicilian-Tunisianchannel in the central Mediterranean has been
shaped by the emergence of two independent
tectonic processes that overlap each other, theaccretionary prism Maghrebides-Apennines and therift Canal Sicilian-Tunisian .
Since at least the Pliocene, these two processes
have acted simultaneously, each being associatedwith subduction Apennines and the African Rift.
Linked to the movement of the accretionary prism,
containing nearly orthogonal normal faults related
rifts and vice versa (Pondrelli, 1999). Analog
modeling supports kinematics inferred from regional
structural data. alkaline magmatism associated withthe fracture is more pronounced in the foreland of
the prism, where the extension is more concentrated.
This particular parameter confirms theindependence of geodynamic processes that caninteract in the same area at the same time suggestingthat the plate boundaries are passive features thatmeet the velocity fields in the far field of thelithosphere (Pondrelli, 1999).

6. DISCUSSION
Analysis of seismic values recorded by local
sources (Institute of Meteorology (Tunisia) andParametric Catalogue of Italian Earthquakes:
CPTI11 (Italy)) are consistent with the AfricanEurasian boundary identified in this work, indicating
a close relationship with plate tectonics associatedwith the movement of the African continent
converge NE-SW in the area of the Sicilian-Tunisian
Strait. The study of the geometry of the seismic
deformation calculated on a larger scale inferredfrom seismicity is a fairly continuous sketch of thedeformation pattern.
Only along the Alps the high seismicity,representative of the regional stress is insufficient toderive a clear tectonic model (Pondrelli 1999). The
heterogeneity of the deformation of this zone has
been confirmed by studies of more small events that
demonstrated the co-existence of contemporaneouscompression and extensional stress regime (Frepoli
and Amato, 1997).
Going North of Tunisia to the island of Sicily,
we see the different characteristics of typicaltranspressive limits that have been shown where asimple compression limit is expected from global
models.
In the central Apennines and southern andeastern Adriatic, the geometry of the seismic
deformation shows that the two areas are directly
involved in the development of peri-Tyrrhenian area
542

GEOMATE-Bangkok, Nov. 14-16, 2016

centro-méridional de Tunisie. 14ème Coll.
Bassins Sédimentaires. Marocains, p. 142.
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A general agreement has estimated that the
Tyrrhenian subduction of the Arc and the South
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Although the study area is part of a broad zone ofdeformation reactivated under compression stressfield current, the study area should suffer the highest
rates of deformation related to the subduction in thesouthern Italy presented by the recent earthquakes ofconsiderable magnitude located in the SicilianTunisian Strait and the Ionian sea and volcanic
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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

REMOVAL OF ORGANIC MATTER IN WASTEWATERS OF A MILKFACTORY AND A HOSPITAL USING A CUBIC LATTICE BASEDROTATING BIOLOGICAL CONTACTOR IN VIETNAM
Tatsuhide Hamasaki1, Phan Do Hung2 and Hiroshi Tsuno3
Faculty of Human Environment, Osaka Sangyo University, Japan;
2
Institute of Environmental Technology, Vietnamese Academy of Science and Technology, Vietnam
1,3

ABSTRACT: Pilot tests of a cubic-lattice-based rotating biological contactor were implemented to remove
organic matter from wastewater from a milk factory and a hospital in Vietnam. In the milk factory wastewater,
the biochemical oxygen demand (BOD) removal ratio was stable between 60% and 90% (average 75%) using
this method, with a BOD surface load of 0.002–0.020 kg•m−2•day−1. The average nutrient ratio of the raw
wastewater was 0.13 of total nitrogen and 0.015 of total phosphorus compared with 1.0 of BOD. The BOD oftreated water was less than 50 mg•L−1, achieving category B of the industrial wastewater standard of Vietnam(QCVN 40:2011). For the hospital wastewater, the BOD removal ratio was stable between 60% and 90%(average 78%), with a BOD surface load of 0.005–0.022 kg•m−2•day−1. The average nutrient ratio of the raw
wastewater was 0.25 of total nitrogen and 0.018 of total phosphorus compared with 1.0 of BOD. The BOD oftreated water was less than 50 mg•L−1, satisfying category B of the medical wastewater standard of Vietnam(QCVN 28:2010/BTNMT). The electric power consumption was 0.73 KWh•m-3 of wastewater. The sludge
conversion ratio from BOD was 0.51 kg TSS•kg BOD−1 based on the excess sludge and suspended solids in raw
wastewater and treated water.
Keywords: Organic Matter, Rotating Biological Contactor, Wastewater, Vietnam
1. INTRODUCTION
A rotating biological contactor comprises anumber of disc-shaped rotating contactors.
Approximately 40% of the contactor surface isdipped into wastewater in a reactor [1]. The
contactors are rotated by an electric motor throughtheir axes (Fig.1). Microorganisms adhere on thesurface of the contactor, forming a film known as a
biofilm. The biofilm decomposes organic matter inwastewater by taking up oxygen when the biofilm
emerges from the wastewater.
Rotating Disc
Raw
Water

Axis

Fig.2 Disk of the rotating biological contactor
This wastewater treatment process does not use ablower for aeration to supply oxygen to wastewater,
and thus does not consume more energy comparedwith the activated sludge process.
This process is widely used for industrial
wastewater treatment in Japan and Europe[3][4][5][6][7], and has been researched anddeveloped for mainly wastewater treatment of small
and medium sized factories, and in developed
countries so far. The manufacturer has experiences
for industrial wastewater treatment using thisprocess, mainly in Japan. However, they are not yet
used in Vietnam and other Asian countries, and nor
are they used for sewage treatment, even in Japan.

Treated
Water

Fig.1 Schematic of the rotating biological contactor
The rotating biological contactor disk used forthis research is illustrated in Fig.2. The disk
comprises a cubic lattice made from polypropylene
to ensure a larger surface area. The manufacturer is
Sekisui Aqua Systems Co., Ltd [2].
545

GEOMATE- Bangkok, Nov. 14-16, 2016

process.
Two pilot tests used the same test plant, thespecifications of which are listed in Table 3.
To confirm the removal capability of organic
matter in terms of BOD and COD, inlet samples
were obtained between the flow control unit and thereactor, and outlet samples were taken between thereactor and the equalizing tank. BOD removal was
calculated using Eq. (1). COD removal was
calculated using Eq. (2) [1].

The manufacturer recently starts manufacturingapparatus of the process in China for promoting
sales.
This research had two aims: to apply this process
to industrial wastewater to remove organic matter in
Vietnam, and to use the process for sewage
treatment to remove organic matter. The sites of thepilot tests of the research were a milk factory and ahospital in Hanoi city.
The main industry of Vietnam is agriculture,mainly rice cropping, with an active livestockindustry in mountainous areas and a substantial
fishing industry on the coast. In recent years, theVietnamese diet has become increasingly
westernized, so the consumption of milk and milk
products has gradually increased and the amount of
milk and milk products in Vietnam has risen.
However, many factories have not conducted
appropriate wastewater treatment, and the water
quality of public waters such as rivers and lakes has
deteriorated. Both factories and public facilities suchas hospitals have not treated wastewater properly.
Industrial wastewater standards have been
established by law (called QCVN 40:2011), as havemedical wastewater standards (called QCVN
28:2010/BTNMT). The values for biochemical
oxygen demand (BOD) and chemical oxygen
demand (COD) of both standards are listed in Tables
1 and 2. BOD analysis uses the five-day method,
based on the Vietnamese standard, as for the Japan
Industrial Standard and USEPA methods. COD
analysis uses potassium dichromate as an oxidizing
agent, based on the Vietnamese standard, which isthe same as the USEPA method.
In the Vietnamese standards, category A meansthat the treated water outlet is located in public waterthat is a water source for domestic and industrialuse; category B means that the treated water outlet is
not located in public water that is a water source fordomestic and industrial use.

Table 3 Specifications of the test plant
Component

Reactor Volume
Diameter of Disk
Total Surface Areaof Disks
Electric Power
Design Drainage
Discharge

Unitmg•L−1mg•L−1Category A
30
75

Unit
mg•L−1mg•L−1Category A
30
50

𝑩𝑶𝑫 𝑰𝒏𝒍𝒆𝒕 !𝑩𝑶𝑫𝑶𝒖𝒕𝒍𝒆𝒕(1)
𝑪𝑶𝑫 𝑹𝒆𝒎𝒐𝒗𝒂𝒍 % =𝑩𝑶𝑫𝑰𝒏𝒍𝒆𝒕

𝑪𝑶𝑫 𝑰𝒏𝒍𝒆𝒕 !𝑪𝑶𝑫𝑶𝒖𝒕𝒍𝒆𝒕(2)

𝑪𝑶𝑫𝑰𝒏𝒍𝒆𝒕

×𝟏𝟎𝟎

×𝟏𝟎𝟎

The BOD surface load was calculated using Eq.
(3) [1].
𝑳𝑺𝒖𝒓𝒇𝒂𝒄𝒆 =

𝑳𝒇 ×𝑸

(3)

𝑨

LSurface: BOD surface load (kg BOD•m−2•day−1)
Lf: BOD inlet (mg•L−1)
Q: Inlet flow (m3•day−1)
A: Total surface area of disks (m2)Category B
50
150

The hydraulic detention time (HDT) was
calculated using Eq. (4) [1].

Table 2 BOD and COD of QCVN 28:2010/BTNMT
Items
BOD
COD

1.5 KW
30 m3•day−1

𝑩𝑶𝑫 𝑹𝒆𝒎𝒐𝒗𝒂𝒍 % =Table 1 BOD and COD of QCVN 40:2011
Items
BOD
COD

Flow Control Unit –Reactor with Disks –Equalizing Tank –Sedimentation Tank
1.78 m3
1.2 m
420 m2

𝑯𝑫𝑻 𝒉𝒐𝒖𝒓𝒔 =Category B
50
100

𝑹𝒆𝒂𝒄𝒕𝒐𝒓 𝑽𝒐𝒍𝒖𝒎𝒆 𝒎𝟑
𝑸 𝒎𝟑 ∙𝒅𝒂𝒚!𝟏

×𝟐𝟒

(4)

Equation (5) was used to calculate the hydraulicloading rate [8].
𝑯𝒚𝒅𝒓𝒂𝒖𝒍𝒊𝒄 𝒍𝒐𝒅𝒊𝒏𝒈 𝒓𝒂𝒕𝒆 𝒎𝟑 ∙ 𝒎!𝟐 ∙ 𝒅𝒂𝒚!𝟏 =2. OBJECTS AND METHODS

𝑸 𝒎𝟑 ∙𝒅𝒂𝒚!𝟏

𝐓𝐨𝐭𝐚𝐥 𝐒𝐮𝐫𝐟𝐚𝐜𝐞 𝐀𝐫𝐞𝐚 𝐨𝐟 𝐃𝐢𝐬𝐜𝐬 𝒎𝟐

Objects of the research are to remove organic
matter such as BOD and COD under each standard
continuously and confirm sludge production of this(5)

In addition, the total nitrogen (TN) and total
phosphorus (TP) levels of the samples were546

GEOMATE- Bangkok, Nov. 14-16, 2016

analyzed to assess the relations of these parameterswith BOD, based on Vietnamese standards.

2.2 Hospital
The hospital is located in Hanoi city. The
hospital has approximately 800 beds and an averagedrainage discharge of 400 m3•day−1. The hospital has
a wastewater treatment plant that uses an anaerobic
treatment process and a trickling filter process. The
wastewater is mainly from the linen room.
The pilot test plant was installed in front of theexisting wastewater treatment plant. During the pilot
test, raw water samples were obtained at theequalizing tank of the existing wastewater treatment
plant, as shown in Fig.5.
The pilot test at the hospital was conducted from
July to December 2014. During this period, samples
were taken at the inlet and outlet on thirteen
occasions. The inlet BOD was 51–270 mg•L−1, inlet
COD was 91–300 mg•L−1 ，inlet TN was 21–44
mg•L−1, and inlet TP was 1.1–3.5 mg•L−1. The
average ratio of BOD, TN, and TP was 100 BOD:25
TN:1.8 TP. The average inlet flow was 31 m3•day−1,
so the average HDT was approximately 1.4 hours
and the hydraulic loading rate was 0.07m3•m−2•day−1.
For the hospital, the production of the sludge
generated in the reactor was calculated.
Figure 6 shows rotating disks with biofilm in thereactor at the hospital.

2.1 Milk Factory
The milk factory is located in Hanoi city. The
designed drainage discharge of the factory is 300
m3•day−1; however, it was operated at 100 to 250
m3•day−1 during the pilot test. The factory has awastewater treatment plant that uses a trickling filter
process and an activated sludge process.
The pilot test plant was installed in front of theexisting wastewater treatment plant. Water samples
in the form of raw water for the pilot test wereobtained at the equalizing tank of the existing
wastewater treatment plant, as shown in Fig.3.
The pilot test at the milk factory ran from July
2013 to January 2014. During this time, samples
were taken at the inlet and outlet on 28 occasions.
The inlet BOD was 40–320 mg•L−1, inlet COD was
140–640 mg•L−1，inlet TN was 1.0–46 mg•L−1, andinlet TP was 0.2–3.3 mg•L−1. The average ratio ofBOD, TN, and TP was 100 BOD: 13 TN: 1.5 TP.
The average inlet flow was 29 m3•day−1, so theaverage HDT was approximately 1.5 hours and thehydraulic loading rate was 0.07 m3•m−2•day−1.
Figure 4 shows rotating disks with biofilm in thereactor at the milk factory.

Fig.5 The pilot test plant at the hospital

Fig.3 The pilot test plant at the milk factory

Fig.6 Rotating disks with biofilm at the hospital
Fig.4 Rotating disks with biofilm at the milk factory547

GEOMATE- Bangkok, Nov. 14-16, 2016

3. RESULTS AND DISCUSSION
3.1 Milk Factory
Figure 7 shows the BOD of the inlet and theoutlet, and the BOD surface load for the milk factory.
The average BOD surface load was 0.009 kg
BOD•m−2•day−1. At the beginning of the test, BOD
removal was lower, but it gradually increased withtime. The average value of BOD removal was 75%.
The BOD outlet value was less than 50 mg•L−1,
falling into category B of the industrial wastewater
standards of Vietnam, except at the beginning of thetest, and the BOD of half of the samples was lessthan 30 mg•L−1, satisfying category A of theVietnamese standard.
Figure 8 shows the relation between BOD
removal and BOD surface load. These results
indicate that more than 60% of BOD could be
removed irrespective of the BOD surface load,
which was less than 0.020 kg BOD•m−2•day−1 inalmost all cases.
Figure 9 shows the COD measurements for the
inlet and outlet of the milk factory. Similarly to the
BOD results, COD removal was lower at the
beginning of the pilot test; however, COD removal
gradually rose, and the average COD removal was
72%. According to the industrial wastewater
standards of Vietnam, 64% of the outlet COD values
were less than 75 mg•L−1, falling into category B.

Fig.8 Relation between BOD removal and BOD
surface load for the milk factoryFig 9. COD of the milk factory
3.2 Hospital
Figure 10 shows the inlet and outlet BOD and
BOD surface load of the hospital. The average BOD
surface load was 0.010 kg BOD•m−2•day−1. The
BOD of the inlet gradually increased from 51
mg•L−1 to 270 mg•L−1; however, the outlet BOD
was always less than 50 mg•L−1, falling into
category B of the medical wastewater standards ofVietnam. Approximately 60% of the outlet BOD
measurements were less than 30 mg•L−1, satisfying
category A of the Vietnamese standard.

Fig.7 BOD and BOD surface load of the milkfactory548

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig.10 BOD and BOD surface load of the hospital

Fig.12 COD of the hospital

Figure 11 shows the relation between BOD
removal and BOD surface load. The graph indicatesthat more than 60% of BOD could be removed inmost cases, irrespective of the BOD surface load,
which was less than 0.015 kg BOD•m−2•day−1.

For the calculation of the sludge production in
the reactor, samples from the inlet and outlet andsamples after the sedimentation tank were obtained
to measure BOD and TSS in January 2015. The
BOD and TSS results are provided in Table 4. The
sludge in the sedimentation tank of the pilot test
plant was produced at a rate of 0.63 m3•day−1.
The amount of sludge generated was 1.17kg•day−1 and the removed BOD volume was 2.3
kg•day−1, so the sludge conversion rate was 0.51 kg
TSS•kg BOD−1.
Table 4 BOD and TSS values for sludge production
Items

Unit

Inlet

Outlet

BOD
TSS

mg•L−1mg•L−1

92
41

28
36

After Sed.
Tank
2,140

During the period of the pilot test, the pilot test
plant treated 6,140 m3 of water and consumed 4,500
KWh of electric power, representing an electric
power consumption of 0.73 KWh• m-3 of wastewater.

Fig.11 Relation between BOD removal and BOD
surface load for the hospital

4. CONCLUSION

Figure 12 shows the inlet and outlet COD
measurements for the hospital. Similarly to the BOD
results, the inlet COD values gradually increased
from 91 mg•L−1 to 300 mg•L−1. However, the outlet
COD was always less than 100 mg•L−1, falling into
category B of the medical wastewater standards ofVietnam. Approximately 70% of the outlet COD
values were less than 50 mg•L−1, satisfying category
A of the Vietnamese standard.

According to [2], the design criteria for rotating
biological contactors are a hydraulic loading rate of
0.08–0.16 m3•m−2•day−1, a surface load of 0.0098–
0.0172 kg BOD•m−2•day−1, HDT of 0.7–1.5 hours,
and BOD in the effluent of 15–30 mg•L−1. For themilk factory, the hydraulic loading rate was 0.07m3•m−2•day−1, the average surface load was 0.009 kg
BOD•m−2•day−1, HDT was 1.5 hours, and theaverage BOD in the effluent was 30 mg•L−1. For thehospital, the hydraulic loading rate was 0.07549

GEOMATE- Bangkok, Nov. 14-16, 2016m3•m−2•day−1, the average surface load was 0.010 kg
BOD•m−2•day−1, HDT was 1.4 hours, and theaverage BOD in the effluent was 25 mg•L−1. Thus,
both pilot tests almost fulfill the design criteria.
According to [3], the required sludge conversion
rate for rotating biological contactors is 0.75–1.0 kg
TSS•kg BOD−1; however, the sludge conversion rateof the pilot test at the hospital was 0.51 kg TSS•kg
BOD−1.
From this pilot testing, the rotating biologicalcontactor formed of a cubic lattice can be applied forthe removal of organic matter from industrial
wastewater and sewage water in Vietnam. In
addition, there were no mechanical troubles during
the pilot test.

[3] Web Site of Japan Protect of Environment
Co.,Ltd(In
Japanese)
http://www.jpexnet.co.jp/products/index.html[4] Arasuna, M., Advanced Wastewater Treatment
Now -Ideal Development of Rotating BiologicalContactor for Wastewater from Small andMedium Sized Factories-, Journal of Global
Environment, Vol.31 、 No.4 、 p.146-147 、
2000 (In Japanese)[5] Fukuda, F., Improvement of Biofilm Methodsof Resource Recycling Type Wastewater
Treatment Technology -Review of Rotating
Biological Contactor-, Journal of Chemical
Apparatus, Vol.43, No.8, pp.29-35, 2001 (InJapanese)[6] Nishidome, K., etc., Wastewater Treatmentusing Rotating Biological Contactor, Journal of
Water and Waste, Vol.43, No.2, pp.125-140,
2001 (In Japanese)[7] Winkler G., etc., Mass Flow Balances of

5. ACKNOWLEDGEMENTS
This research had been supported by Sekisui
Aqua Systems Co., Ltd and Osaka prefectural
government, and funded by the ministry of theenvironment in Japan.

Triclosan in Small Rural Wastewater Treatment
Plants and the Impact of Biomass Parameters on
the Removal, English Life Science, Vol.7, No.1,
pp.42-51, 2007[8] Marcos von Sperling, Activated Sludge andAerobic Biofilm Reactors. Biological
Wastewater Treatment Series, IWA, Vo1.5,
2007, pp.287-293.

6. REFERENCES[1] “Technology and Regulation for Pollution
Control”, Water Quality Series, Japan
Environmental Management Association for
Industry, 2016 (In Japanese).
[2] Web Site of Sekisui Aqua Systems Co., Ltd
http://www.sekisuia.co.jp/english/plant/product/
esrotae/index.html

550

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

GREEN INNOVATION OF CALCIUM SULFATE OR GYPSUM
PREPARATION FROM DUCK EGGSHELL VIA PYROLYSIS
Nuchnapa Tangboriboon1* and Wanitcha Unjan 2* Materials Engineering Department, Faculty of Engineering, Kasetsart University, Bangkok 10900, Thailand

1 ,2

ABSTRACTCalcium sulfate dihydrate is an important material for construction i.e. wall, roof, and decorative parts.
It can be made from calcium carbonate from duck eggshell react to sulfuric acid at room temperature (25ºC).
Calcium sulfate dihydrate was dried in an oven at 110ºC and calcined at 700°, 800°, and 900°C for 2 hr. It canchange the microstructure from calcium sulfate hemihydrate or plaster of Paris (CaSO 4 .0.5H 2 O) to anhydrite or
anhydrous calcium sulfate. The best condition to obtain the calcium sulfate or gypsum is calcined at 900°C for 2
hr. The best anhydrous calcium sulfate or anhydrite calcined at 900°C for 2 hr has true density, color, specificsurface area, avg. pore diameter, and avg. particle size equal to 2.95 g/cm3, white powder, 3.57 m2/g, 96.98 Å,
and 3.983 µm, respectively. In addition, characteristics, microstructures, phase transformation, and physicalproperties of raw materials and calcium sulfate compounds are reported here by using XRF, SEM, XRD,
pycnometer method, and BET.
Keywords: Calcium sulfate, Duck eggshell, Pyrolysis, Anhydrite, Construction.

differences. The calcium sulfate compounds can beprepared from a variety of processes i.e. dissolved inaqueous sulfuric acid, flue gas desulfurization(FGD),
reversemicro-emulsion,
chemicalprecipitation, ion exchange, adsorption, reverse
osmosis and electro-chemical methods, etc. [13-21].
Eggshell is a calcium carbonate source obtainedby consuming and industries as food, drug,
cosmetics, filler in other industries, etc. The byproduct eggshell represents about 11% of the total
weight approximately 60 g of an egg [22-24]. The
main composition of eggshells is calcium carbonate(CaCO 3 ) known as calcite more than 94 wt% andthe other oxides including organic compounds 6
wt% [22-24]. Eggshell is potential to be used as a
useful waste in a variety of applications [22-27].
The objective in this research, the raw duckeggshell was used as a calcium carbonate to reactwith sulfuric acid under calcination temperature inorder to obtain the calcium sulfate compounds i.e.
calcium sulfate dihydrate or gypsum, hemihydrate orplaster of Paris, and anhydrite. The characteristics,
micro- structures, phase transformation, and physicalproperties of samples were reported here by using
XRF, SEM, XRD, pycnometer method, and BET.

INTRODUCTION
Calcium sulfate is a ceramic material useful formany kinds of industrial applications i.e. a binder forbuilding material, bone graft materials, periodontal
disease treatment, endodontic lesions, alveolar boneloss, maxillary sinus augmentation, filler for plastic,
rubber, coating, construction materials, desiccant,coagulant, and catalysts [1-3]. Calcium sulfate
compounds in terms of γ-or β-anhydrite namely
anhydrous calcium sulfate, calcium sulfate
hemihydrate or plaster of Paris, and gypsum are themost abundant sulfate mineral found in the nature[4]. Natural gypsum or calcium sulfate dihydrate is amonoclinic-prismatic mineral with a layered crystal
lattice containing the water. When the calcium
sulfate dihydrate heated, it can be converted to themetastable hemihydrate and anhydrite type III.
Anhydrite type III can be transformed to anhydrite
type II with the most densely packed ion lattice anddoes not react very readily with water at hightemperature [5]. Therefore, phase transformation,
crystal structure, morphology, and growth rate of
calcium sulfate depend on temperature, pressure,dissolved electrolytes or organics, and otherminerals [4]. Furthermore, there are many types
calcium sulfate i.e. calcium sulfate dihydrate or
gypsum,subhydrate
calcium
sulfate(CaSO 4 .0.81H 2 O), ɑ- and β-hemihydrate calcium
sulfate (CaSO 4 .1/2 H 2 O), calcium sulfate anhydrite
I, II, and III [6-12]. The synonyms, characteristics(color, odor, and density values), crystal structures,chemical solubility performances, physical-opticalmechanical properties (translucence, refractive index,
hardness, etc.) and their applications are some

EXPERIMENTAL
Materials and Methods
Duck eggshell was collected from the cafeteria at
Kasetsart University, Thailand. The duck eggshells
were cleaned with tap water, dried in the air for 2days, and ground with high speed mill for 120 min.
Hydrosulfuric acid (H 2 SO 4 ) is high purity 98% and551

GEOMATE- Bangkok, Nov. 14-16, 2016

was purchased from Arsom Co., Ltd., Thailand.
Hydrosulphuric acid is colorless, odorless, melting
temperature at 104°C and 1.0 atm.

When the CaCO 3 reacts to H 2 SO 4 , the precipitatedcalcium sulfate or calcium sulfate dihydrate powder
was filtered, rinsed with tap water 2-3 times, anddried in the oven at 110°C for 24 hr. Calcium sulfate
dihydrate can transform to calcium sulfate
hemihydrate according to the equation (3). After that,the dried calcium sulfate hemihydrate powder was
calcined at 700°, 800°, and 900°C for 2 hr. Calcium
sulfate hemihydrate converted to anhydrite or
anhydrous calcium sulfate according to the equation(4). The obtained calcined powder was measured thephysical properties and characterized by SEM, XRD,
particle size distribution, BET, and pycnometermeasure ment.

Instruments
Muffle furnace (Nebertherm, Ceramotherm withthermocouple type K, NiCr-Ni) was used to calcine
samples with a heating rate of 10°C/min. The
precipitate calcium sulfate powder prepared from thereaction between duck eggshell and concentrated
sulfuric acid at room temperature was calcined attemperature 700°, 800°, and 900°C for 2 hr with aheating rate of 10ºC/min. High speed mill model
RM 1105 with speed 500 rpm was supplied byCompound Clay Co., Ltd., Thailand. The rapid mill
is a porcelain pot containing the amount of 2/3
porcelain ball mills of the porcelain pot volume. The
rapid mill was used for grinding the duck eggshell to
be fine powder for the calcium sulfate preparation.
X-ray diffraction (XRD) was taken and analyzed
using a Bruker AXS analyzer (D8 Discover) with
VANTEC-1 Detector. Samples were analyzed using
a double-crystal wide-angle goniometry. Scans were
measured from 5°- 80° 2θ at a scan speed of 5°
2θ/min in 0.05° or 0.03° 2θ increments using CuK α
radiation (λ = 0.15406 nm). Peak positions wereconsistent with those of the International Center for
Diffraction Data Standard (JCPDS) patterns to
identify crystalline phases. Scanning Electron
Microscope (SEM) was taken and characterizedusing SEM, JEOL-5200. The samples of raw
material and calcium sulfate compound were
mounted on a stub using carbon paste and weresputter-coated to ~0.1 µm of gold to improve
conductivity. The acceleration voltages of 11 and 13
kV with magnifications of 1000 and 5000 times
were used.
Calcium Sulfate Powder Preparation (CaSO 4 .
2H 2 O, CaSO 4 .0.5H 2 O, and CaSO 4 )RESULTS AND DISCUSSION
Characteristics of Raw Material and Calcium
Sulfate Powder
The chemical composition comparison of duckeggshell and calcined duck eggshell was measured
by XRF as data tabulated in Table 1. The main raw
duck eggshell composed of calcium carbonate(CaCO 3 ) 98.101 wt% and other oxide compounds
1.899 wt%. While the calcined duck eggshell
composed of calcium oxide (CaO) 97.805 wt% andthe other oxide compounds such as MgO, Na 2 O,
K 2 O, SiO 2 , etc. 2.195 wt%. However, both of raw
duck eggshell and calcined duck eggshell can react
to sulfuric acid at room temperature (25ºC)
according to the equations (1) and (2) to obtaincalcium sulfate dihydrate or gypsum. When calcium
sulfate dihydrate was heated, it can be transformed
to calcium hemi hydrate or plaster of Paris andanhydrite according to equations (3) and (4),respectively.
Physical Properties and Microstructures ofCalcium Sulfate Powder

The duck eggshell ground for 120 min acted as
calcium carbonate source (CaCO 3 ) to react withsulfuric acid at room temperature (25ºC) as
equations (1), (2), (3) and (4):2CaCO 3 +2H 2 SO 4
(1)

The physical properties (average particle size,true density, specific surface area, and average pore
diameter) of raw duck eggshell and anhydrous
calcium sulfate calcined at 800º and 900ºC for 2 hr
were measured and reported as data tabulated inTable 2. The particle size distribution at d 90 , d 50 , d 10 ,and d avg of raw duck eggshell is 90.02, 15.75, 1.55,
and 34.35 µm, respectively. The true density,
specific surface area, and average pore diameter ofthe raw duck eggshell are 2.25 g/cm3, 7.79 m2/g, and
196.90 Å, respectively. While the particle size
distribution at d 90 , d 50 , d 10 , and d avg of calcium
sulfate calcined at 900°C for 2 hr is equal to 7.39,
3.13, 1.08, and 3.99 µm, respectively. Furthermore,
the true density, specific surface area, and averagepore diameter of calcium sulfate calcined at 900°C
for 2 hr for 2 hr are 2.95 g/cm3, 3.57 m2 / g, and 96.98

2CaSO 4 .2H 2 O+2CO 2

Gypsum or calcium dihydrate
2CaO+2H 2 SO 4
(2)
2CaSO 4 .2H 2 O(3)

2CaSO 4 .2H 2 O
CaSO 4 .0.5H 2 O+1.5H 2 O

Calcium hemihydrate
CaSO 4 .0.5 H 2 O(4)CaSO 4 +0.5H 2 O

552

GEOMATE- Bangkok, Nov. 14-16, 2016

Å, respectively. The porosity of calcium sulfate
calcined at 900°C for 2 hr is in the range of mesoporous structure (20Å-500Å). The physicalproperties of the raw duck eggshell i.e. particle sizeand shape, specific surface area, true density,
solubility, etc., are important factors of calcium
sulfate formation and its applications [6, 29-31].
Calcined duck eggshells

Raw duck eggshells
Compounds
Na 2 O
MgO
Al 2 O 3
SiO 2
P2O5
SO 3
Cl
K2O
CaCO 3CuO
SrO

32.204°. The main XRD peak pattern of the driedprecipitated calcium sulfate powder before firingshows the rhombohedral structure of calcium sulfate
hemihydrate consistent with JCPDS file no. 01-0700909 at the (hkl): (020) 25.432º, (104) 29.364º,(012) 31.366º, (022) 38.648º, (212) 40.820º, and(032) 48.696º. Then, the dried precipitated samples,
calcium sulfate hemihydrate powder calcined at
700° and 800°C for 2 hr, show the same XRD peak
patterns of metastable phase of anhydrous calcium
sulfate in hexagonal phase formation consistent with
the JCPDS file nos. 01-089-1458 and 01-070-0909
at the (hkl): (100) 14.665°, (200) 29.577°, (102)32.011° and (020) 25.432°, (012) 31.366°, and (212)40.820°, respectively, mixed with small amount of
the calcium hydroxide or portlandite (Ca(OH) 2 )belong to orthorhombic consistent with the JCPDS
file no. 00-004-0733 at the (hkl): (101) 34.089°,(102) 47.124°, and (110) 50.795°. Furthermore, thedried precipitated calcium sulfate dihydrate sample
calcined at 900°C for 2 hr, shows the stable
crystalline phase formation of hexagonal namely
anhydrite structure (CaSO 4 ) consistent with the
JCPDS file no.01-089-1458 at the (hkl): (020)25.432°, (012) 31.366°, and (022) 38.648°,
respectively, consistent with the anhydrite crystal
structure obtained by Zhao, Wu et al [3]. Therefore,
the obtained XRD phase transformation of calcium
sulfate dihydrate suggests it can transform tocalcium sulfate hemihydrate or plaster of Paris(CaSO 4 .0.5H 2 O) or anhydrite crystal structures
depending on firing temperature and firing time
consistent with the calcium sulfates equations (1)-(4)of dehydration-rehydration.
The SEM micrographs of duck eggshell powder,calcium sulfate hemihydrate powder before firing,and dried precipitated calcium sulfate hemihydrate
powder calcined at 700°, 800°, and 900°C with themagnifications of 1000 and 5000 times as shown in
Fig. 2. The SEM micrographs of ground raw duckeggshell powder show the particulate agglomerationand non-uniform size as shown in Figures 2a and 2a1. The SEM micrographs of the precipitated calcium
sulfate hemihydrate powder dried at 110ºC without
firing show uniform needle shape crystal structure
consistent with the results obtained by Freyer, D etal. [4], Licong, D. et al. [13], and Azimi G. et al.
[32]. The SEM microstructures show particles
agglomeration and microstructure changing from theneedle shape to plate- or disk-like shape due tophase transformation at the firing temperature 700°C
as shown in Figures 2c and 2c-1 consistent with theSEM results obtained by Gartner, E.M. [33]. Then,the microstructures changed to more plate-like due
to crystallization as a metastable phase at firing
temperature 800°C as shown in Figures 2d and 2d-1.
Furthermore, the microstructures changed andcrystallized completely to form the small rod-like inshape belonging to stable anhydrite at calcination

Weight
(%)
0.204
0.286
0.035
0.073
0.443
0.764
0.035
0.038
98.101
0.009
0.013

Compounds
Na 2 O
MgO
SiO 2
P2O5
SO 3
Cl
K2O
CaO
SrO

Weight
(%)
0.161
0.656
0.255
0.775
0.215
0.053
0.059
97.8050.019

Table 1 Chemical composition of raw materials
measured by XRF
SamplesRaw
duckeggshell
CaSO 4
calcined
at 800ºC
CaSO 4
calcined
at 900ºC

Avg.
particle
size
(µm)

True
density(g/cm3)Specificsurfacearea(m2/g)

Avg.
pore
diameter
(Å)

34.35

2.25

7.79

196.90

5.560

2.87

N/A

N/A

3.983

2.95

3.57

96.98

Table 2 Physical properties of samples

The XRD peak patterns of raw duck eggshell
before firing and dried calcium sulfate compounds
obtained from raw duck eggshell react to sulfuric
acid calcined at 700°, 800°, and 900°C, for 2 hr, as
shown in Fig. 1. The XRD peak pattern of raw duckeggshell shows the crystalline phase formation ofrhombohedral or calcite consistent with the JCPDS
file no. 01-086-2339 at the (hkl): (104) 29.364°,(012) 23.058°, and (113) 39.424° while the XRD
peak pattern of calcined raw duck eggshell shows
the crystalline phase formation of lime or calcia
consistent with the JCPDS file no. 00-037-1497 atthe (hkl): (200) 37.347°, (220) 53.856°, and (111)553

GEOMATE- Bangkok, Nov. 14-16, 2016

temperature 900°C as shown in Figures 2e and 2e-1
consistent with the SEM results obtained by Azimi
G. et al. [32].

Calcium sulfate 900°C

Calcium sulfate 800°C

c)

c-1)

d)

d-1)

e)

e-1)

Calcium sulfate 700°C

Calcium sulfate hemihydrate
before firing

Eggshell fired at 900°C

Raw duck eggshells

Fig.1 XRD peak patterns of raw duck eggshell
beforecalcination and dried calcium sulfate
hemihydrate powder calcined at 700°, 800°, and
900°C, for 2 hr, respectively.

Fig. 2 SEM micrographs of samples with themagnifications of 1000 and 5000 times: a) and a-1)
raw duck eggshell powder; b) and b-1) calcium
sulfate hemihydrate (CaSO 4 .0.5H 2 O) powder before
firing; c) and c-1) calcium sulfate hemihydrate
powder calcined at 700°C; d) and d-1) calcium
sulfate hemihydrate powder calcined at 800°C; ande) and e-1) calcium sulfate hemihydrate powder

a)

a-1)

calcined at 900°C.
CONCLUSIONS

b)

b-1)

Duck eggshell is a potential starting material for
use as a calcium carbonate (CaCO 3 ) source to reactwith sulfuric acid in order to prepare the calcium
sulfate dihydrate or gypsum, calcium sulfate
hemihydrate or plaster of Paris, and stable anhydritein terms of CaSO 4 .xH 2 O (x = 0.0-2.0) by thermal

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GEOMATE- Bangkok, Nov. 14-16, 2016

Chulalongkorn University, the Thailand Research
Fund (TRF), and the Royal Thai Government.

process. Calcium sulfate can form hydration anddehydration process due to water adsorptiondesorption ability within the layer microstructure.
Calcium sulfates have a potential candidate
functioned as a binder, filler, adsorbent, catalyst, andcoagulant in a variety of buildings, ceramics,
petroleum and petrochemical, dental and mechanicalindustries. Calcium sulfate dihydrate or gypsum is
one of important materials suitable for building andmold making, etc., whereas anhydrite or anhydrous
calcium sulfate is suitable for function as a filler invarious industries such as paint, plastic, rubber,coating, cement, etc. [34-36]. There are many
advantages of chemical precipitation method used inthis study for calcium sulfate compounds
preparation i.e. easily and convenient forming, low
price, and high purity calcium sulfate including
waste eggshell reduction. The obtained calcium
sulfate dihydrate or gypsum can form at room
temperature. When the calcium sulfate dihydrate
was dried at 110ºC and calcined at 700°, 800°, and
900°C for 2 hr effect to the microstructure changing.
The microstructure of calcium sulfate dihydrate cantransform to calcium sulfate hemihydrate or plasterof Paris and then transform to anhydrite or
anhydrous calcium sulfate (CaSO 4 ) type III, type II,
and type I. The XRD phase formation of calcium
sulfates will change from rhombohedral (gypsum) to
hexagonal (anhydrite) crystal structure. The SEM
micrograph will change from needle-like shape(hemihydrate or plaster of Paris) to plate- or disklike, and finally to rod-like shape (anhydrite). The
suitable condition to receive the completely
anhydrous calcium sulfate is firing at 900°C for 2 hr.
The true density, color, odor, specific surface area,
average pore diameter, and average particle size ofanhydrous calcium sulfate calcined at 900ºC for 2 hr
are equal to 2.95 g/cm3, white powder, odorless,
3.57m2/g, 96.98 Å, and 3.983 µm, respectively.

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ACKNOWLEDGEMENTS
The authors sincerely thank the Petroleum andPetrochemical College, and the Scientific andTechnological Research Equipment Center, atChulalongkorn University, Thailand; the Departmentof Materials Engineering, at Kasetsart University forthe use of analytical instruments. We have to thankthe Energy and Environmental Engineering Center,
Faculty of Engineering, Kasetsart University for
eggshell classification. This work is grant supported
from Agricultural Research Development Agency(Public Organization) or ARDA of Ministry ofAgricultural and Cooperatives, the Thailand
Research Fund, encoded PRP5605010080. In
addition, we also would like to acknowledge thefinancial supports from the Conductive andElectroactive Polymers
Research Unit of555

GEOMATE- Bangkok, Nov. 14-16, 2016

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[24] Murakami, F. S. and Rodrigues, P.O., Physicochemical study of CaCO 3 from egg shells,Ciénc. Technol. Aliment., Campinas, Vol. 27/3,
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Technol., Vol. 58/1, 2011, pp. 33-41.
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sapong,R., and Sirivat,A., REMOVINGEXTRACTABLE PROTEINS IN NATURAL
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CHLORIDE FROM CHICKEN EGGSHELLS,RUBBER CHEM. TECHNOL., Vol. 84/4,
2011, pp. 543-564.
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sulphate hemihydrate hydration leading to
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Bonnaud, L., and Dubois, P., Polylactide(PLA) and highly filled PLA-calcium sulfate
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Degrad. Stab., Vol. 96, 2011, pp. 616-623.

556

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

VELOCITY STRUCTURE AND EARTHQUAKE RELOCATIONS AT
CENTRAL PENINSULAR MALAYSIA REGION
Abdul Halim Abdul Latiff1 and Amin Esmail Khalil2
1

Faculty of Geoscience and Petroleum Engineering, Universiti Teknologi PETRONAS, Malaysia
2
School of Physics, Universiti Sains Malaysia, Malaysia
ABSTRACTIn between 2007 to 2009, there are several earthquake occurrences in Bukit Tinggi and Janda Baik, which arelocated 50 km from Malaysia capital, Kuala Lumpur. However, the absent of subsurface structural informationand crustal velocity data for central peninsular Malaysia region had prevent further analysis on the potential
seismic hazard in the area. Thus, this study was carried out in determining the new velocity structure of central
Peninsular Malaysia region, while relocate the hypocenter of the local earthquake. Initially, a data conditioning
procedure was carried out for both weak and strong motion data from various sources. Then, P-wave and S-wave
arrivals were picked on the waveform recorded, before we derived a 1-D velocity model through simultaneous
inversion process. The process involved also produced a new hypocenter location based on the new 1-D velocity
structure, correcting the original earthquake location. The outcome from the data analysis was used to understandthe subsurface images and tectonics system of the central peninsular Malaysia region while overcome theuncertainty surrounding the earthquakes occurrences in the area. In addition, the study also contributes to updating
current geological and geophysical map as well as tectonic regime of Peninsular Malaysia while the 1-D velocity
model will be used as an initial reference model for further analysis in 3-D tomography inversion.
Keywords: Earthquake, 1-D Model, Peninsular Malaysia
INTRODUCTION

This geological model of Peninsular Malaysia, interm of crustal thickness is directly correlated to themagnitude and depth of localized earthquakes [3].
With this in mind, there is an urgency to develop the
local velocity structure before further analysis can be
carried out.

Although Peninsular Malaysia lies on the stable
Sundaland and generally safe from major earthquake
disaster (5.0 Mw and above), the small scale tremorsthat occurred in Bukit Tinggi, Pahang in between
2007 to 2009, might give a cause of concern [1]. Up
till today, there is no proof that these seismic
activities are stable and will not strike again in thenear future. It was deduced that the short-time frame
but frequent small scale tremors could have occurred
due to re-activation of paleo-faults lines that passthrough the Peninsular Malaysia, such as Bukit
Tinggi and Kuala Lumpur fault lines (Fig. 1).
However, due to the limitation of Peninsular
Malaysia seismological data analysis and its associate
information, previous research unable to confirm thecurrent state of fault line movements.
Without proper study and research work on
subsurface imaging, these small earthquakes which
release an equivalent 15,000 tons of TnT energy,
might re-occur in the future with a larger amount ofenergy and strength. The existing magnetic andgravity image data of Peninsular Malaysia subsurface
might able to reveal the basinal structures of thesezones, but were less accurate compare to available
seismic methodology which is very expensive. In
recent years, there are various theory relating thepresent day Peninsular Malaysia structure to thecollision of Sibumasu Terrane and Sukhothai Arc [2].

Fig. 1 The seismotectonic of Peninsular Malaysia,
showing the Bukit Tinggi and Kuala Lumpur fault
lines.
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The Bukit Tinggi and Kuala Lumpur fault lines
are the two paleo-fault lines that are said to be in theprocess of re-adjustment in order to accommodate thebuild-up tectonic pressures originating from Indo–
Eurasian plates movements [4]. Generally, theresultant faults re-activation can be proven through acomprehensive 3-D velocity tomography withmagnetic, gravity and geological data correlation. In
the last ten years, seismologist discovered that thereliable subsurface information can also be obtainedby correlating two or more seismometers data, which
will produce a result as if there had been a realearthquake or seismic activity at the otherseismometers [5]. This technique, called seismic
interferometry is being developed further byincorporating ambient noise signal to produce thetomography update [6].

structure as well as the existing hypocenters. In thisprocedure, which commonly implemented using
VELEST software [7], the calculation of a minimum
1-D model requires multiple iterations with selected
control parameters, in order to achieve a uniquesolution. On the other hand, the receiver function
method is a phase conversion technique, which
requires P and S-waves arrivals from the teleseismic
data. The receiver information then can be derived bydeconvolving the horizontal component from verticalcomponent, thus reveal the structure for crustal andupper mantle [8].
DATA DESCRIPTION
Data used for travel time inversion was obtained
from Malaysia Seismological Network (MSN), which
is being operated by Malaysia Meteorological
Department (MMD), a government agency under
Ministry of Science, Technology and Innovation(MOSTI). At the moment, there are 44 seismology
stations being used within Peninsular Malaysia,
Sabah and Sarawak, where 17 of them are the weak
motion stations. However, a number of stations were
only installed after the earthquake occurrences in
Bukit Tinggi. In order to enhance the recorded
seismological data in the affected area, three of thestations were installed in Bukit Tinggi, Goh Tong
Jaya and Janda Baik (Fig 3).

However, before we carry out the 3-D velocity
tomography of the area studied a 1-D priori velocity
model need to be established. While there are anumber of global velocity model available and beenused extensively such as IASP91 (Fig. 2) and AK135
models, the local crustal models are often been
ignored. This local information is essential inprocessing the earthquake data, developing crustal
structure and locating the fault lines. Therefore, a few
local velocity model determination techniques being
experimented, among them are inversion of arrival
times and receiver function method.

Fig. 2 IASP91 model that was constructed as areference and parametrizes velocity model.
Fig. 3 The location of ten earthquake in Bukit
Tinggi (blue pins), and the nearest seismology
stations (yellow pins).

The inversion of arrival times works by inverting
the well-located hypocenters data to improve the local

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Table 1 The list of earthquake occurrences within Peninsular Malaysia in between 2007 and 2013.
No.
1
2
3
4
5
6
7
8
9
10
11
12
13
14
15
16
17
18
19
20
21
22
23
24
25
26
27
28
29
30
31

Date
2013-08-20
2009-12-04
2009-11-30
2009-11-30
2009-11-29
2009-11-29
2009-10-08
2009-10-07
2009-10-07
2009-10-07
2009-10-07
2009-10-07
2009-04-29
2009-03-27
2008-05-25
2008-03-15
2008-03-14
2008-03-14
2008-01-14
2008-01-13
2008-01-13
2008-01-10
2007-12-31
2007-12-12
2007-12-09
2007-12-06
2007-12-04
2007-12-04
2007-11-30
2007-11-30
2007-11-30

Time (UTC)00:26:27
01:41:45
06:29:48
01:12:30
16:15:05
06:26:51
04:05:55
22:20:59
22:09:45
21:51:11
21:26:05
21:21:26
13:53:54
01:46:25
01:36:22
00:50:57
23:35:34
23:16:18
15:45:00
10:18:00
02:24:00
15:38:00
09:19:00
10:01:00
12:55:00
15:23:00
19:57:00
10:12:00
12:42:00
02:42:00
02:13:00

Latitude5.41603.37262.73102.73822.73632.73983.27003.34383.30303.35383.38903.34954.15003.86213.36003.33003.30003.33003.42003.33003.31003.39003.32003.47003.33003.36003.37003.36003.31003.34003.3600Longitude
101.3600
101.8038
102.0670
102.1432
102.1169
102.0910
101.8270
101.8135
101.8340
101.8218
101.9020
101.8094
100.7290
102.5194101.7500
101.7100
101.8600
101.7400
101.8000
101.8300
101.8300
101.7300
101.8100
101.7600
101.8200
101.8100
101.8000
101.8100
101.8400
101.8000
101.8000

Magnitude
4.10
1.95
3.50
3.04
3.30
3.10
0.98
0.31
3.16
4.23
1.02
1.66
2.76
3.24
2.60
3.30
2.50
2.90
3.40
2.40
2.50
3.00
2.60
3.20
3.50
2.70
3.30
3.00
3.20
2.80
3.50

Depth
1.60
5.04
14.67
4.00
1.75
3.00
10.00
1.89
10.00
3.00
10.00
1.94
22.55
50.00
Shallow
Shallow
Shallow
Shallow
Shallow
Shallow
Shallow
3.00
Shallow
Shallow
4.90
Shallow
Shallow
Shallow
6.70
Shallow
2.30

Earthquake Location
Tasik Temenggor
Bukit Tinggi
Kuala Pilah
Kuala Pilah
Kuala Pilah
Kuala Pilah
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Manjung
Jerantut
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi
Bukit Tinggi

Fig. 4 The waveforms for earthquake on 4th December 2009 which was recorded by stations in Peninsular
Malaysia.
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For this study, we incorporate waveform timeseries and arrival times from local earthquakes in
Peninsular Malaysia in between November 2007 to
August 2013 (Table 1). In the analysis conducted, we
manually picked the time arrivals for both the P and
S phases in all 31 earthquakes. After several roundsof phase picking and analysis, it was decided todiscard 14 of the events due to poor data condition as
the main phase information has been masked with
ambient noise signal. In this initial observation, it was
found that the poor waveform data is due tounavailability of weak motion data (not installed atthat moment) which should be able to detect weak
earthquakes below 4.0 Mw. Further analysis theneliminate another 8 earthquakes that were not locatedin the central Peninsular Malaysia, as our objective is
to look into velocity structure in central peninsular
Malaysia region.

determining 1-D velocity model of this region. The
simultaneous inversion approaches provide three
outputs; updated hypocentral parameters, crustal
model and station corrections. In addition, it should
be note that a single iteration of simultaneous
inversion is not sufficient, as the model produced is
not minimum.

METHODOLOGY
To obtain the minimum local 1-D velocity model,
several steps beforehand are needed. Once the data
was acquired and reformatted,
waveform
conditioning and processing such as filtering andphase picking were conducted, by using the SEISAN
software [9]. In the phase picking, only the clear P or
S arrivals will be chosen, while the ambiguous
arrivals will be flag off. Throughout the process,several earthquakes were discarded due to data
insufficiency. All the events in the list are local
earthquake, thus it is natural that the station close to
the event recorded clear phase arrival compare to theone further away from the event (Fig. 4). Although
the phase can also be read as P g and Sg, we picked
them as P and S-wave respectively. With regards tothe area under study, the identification of theminimum 1-D model has been carried out based on
34 P-observations and 14 S-observations. All Sphases were picked on the horizontal component
stations which explain the smaller number of picks.

Fig. 5 The workflow used in determining the
updated 1-D model of central Peninsular Malaysia
DATA INVERSION
The inversion was run until the earthquake
locations, station delays and velocity values did notvary significantly in the subsequent iterations. From
the resultant P-wave velocity model obtained (Fig. 6)
it shows a significant velocity reduction of 4.5% at
the Moho boundary of 31 km depth. By assuming thecrustal depth is constant for both models, the updated
1-D model indicate that central Peninsular Malaysia
region contain a slightly different material compare to
regional velocity of the world. Meanwhile, the Swave velocity (Fig. 7) shows a slightly faster velocity
at the Moho depth, compare to the input model. Thesefindings again reinforce our belief that the central
Peninsular Malaysia was made up from slightly
different granitic block compare to surrounding
region. Another importance from this study wasdescribed by the earthquake relocations (Table 2).
From the outcome of simultaneous inversion, there
are slight differences in the origin time of theearthquake, its longitude and latitude as well as thedepth of the seismic events. Although the changes areminimal, the new hypocenter parameters can be used
for future determination of 3-D velocity tomography
analysis in the region.

In next step, the crustal model was determined bysimultaneously inverting the priori waves’ arrival,
initial 1-D models of both Vp and Vs, as well as station
coordinates (Fig. 5). In the 1-D minimum model, thevelocity of a given layer is considered as the best
average lateral velocity that belongs to that layer.
Since the precision of routine earthquake location isstrictly linked to the accuracy velocity model being
used, we had chosen IASP91 velocity model as theinitial velocity model for this region. In addition, as
our target velocity model of crustal area is in central
Peninsular Malaysia region, we decided to discard
other seismic events that located in other Peninsular
Malaysia region, such as in Manjung and Temenggor,
in the simultaneous inversion procedure. Eventually,
there are only nine events that were used in
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Table 2 The hypocenter relocation parameter of 9earthquakes in Peninsular Malaysia.
Time
Latitude Longitude
Depth
Differences Differences Differences Differences
4th December 2009 0141
4.9
-0.029-0.1280.4
8th October 2009 0405
2.8
0.024
-0.011
0.5
4th December 2007 1011
-0.1
0
0.004
2.5
7th October 2009 2149
2.8
0.021
-0.013
0.7
7th October 2009 2121
3.5
-0.0420.002-2.3
12th December 2007 1001
-0.6
-0.003-0.0160
9th December 2007 1255
0.1
0.003
0.014
4.3
30th November 2007 1242
-0.5
0.001
-0.011
0.1
30th November 2007 0213
-0.2
0.001
0.007
1.3
Date

Time

CONCLUSIONS
Although the results shown are encouraging, caution
must be exercised when interpreting the 1-D velocity
model as no other geological and geophysical models
prove otherwise. The updated 1-D velocity models
has shown that the Peninsular Malaysia has a slightly
different subsurface characteristic compare to othercontinental velocity structure. The application ofsimultaneous inversion techniques to a small scaleregion, where the seismic networks are fully
operational can improve the quality of the 1-D model,
hypocenter parameter relocations and enhance station
correction. It is suggested that the 1-D model
produced is to be used as the initial 3-D velocity
tomography model process.

Fig. 6 1-D VP model for Peninsular Malaysia
before and after iterations.

ACKNOWLEDGEMENTS
We are grateful to Universiti Teknologi PETRONAS
for the opportunity to conduct this research, Malaysia
Meteorological Department for the earthquake
waveforms and to Ministry of Higher Education,
Malaysia for the financial grant support.
REFERENCES[1] Malaysian Meteorological Department (MMD),
2008.http://www.kjc.gov.my/english/service/sei
smology/seismoevent.html.
[2] Metcalfe, I., (2013). Tectonic evolution of the
Malay Peninsula, Journal of Asian Earth
Sciences 76, pp.193-pp213.
[3] Hagen, M., (2016). Induced earthquakes
correlations with earth’s crustal thickness,
International Journal of GEOMATE, Vol. 11,
Issue 23, pp.2228 - pp.2234.
[4] Shuib, M.K. (2009). The recent Bukit Tinggi
earthquakes and their relationship to major

Fig. 7 1-D VS model for Peninsular Malaysia
before and after iterations.

561

GEOMATE- Bangkok, Nov. 14-16, 2016

geological structures, Bulletin of the Geological[7] Kissling, E., Elssworth, W.L., Eberhart-Phillips,
D., Kradolfer, U., (1994). Initial reference
models in seismic tomography, Journal ofGeophysical Research, Vol. 99, No. B10, pp
19635 – 19646.
[8] Ammon, C. J. (1991). The isolation of receiver
effects from teleseismic P waveforms. Bull.
Seism. Soc. Am. Vol. 81, pp. 2504–2510.
[9] Ottemoller, V. and Havskov, (2014). Seisan
Eearthquake Analysis Software for Windows,
Solaris, Linux and Macosx.

Society of Malaysia 55, pp. 67 – 72.

[5] Wapenaar, K., Draganov, D., Snieder, R.,
Campman, X., and Verdel, A., (2010). Tutorial
on seismic interferometry: Part 1 — Basicprinciples and applications, Geophysics, Vol. 75,
No.5, pp.. 195 - 209.
[6] Nicolson, H., Curtis, A., Baptie, B., Galetti, E.,(2012). Seismic interferometry and ambient
noise tomography in the British Isles,
Proceedings of the Geologists’ Association, Vol.
123, pp. 74–86.

562

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

VEGETATIVE INFLUENCE ON ROUGHNESS LEVELS FOR
PAVEMENTS FOUNDED ON ALLUVIAL EXPANSIVE SOILDEPOSITS
Md Yeasin Ahmed1, Robert Evans1 and Monzur Imteaz1
Department of Civil and Construction Engineering, Faculty of Science Engineering and Technology,
Swinburne University of Technology, Melbourne, Australia

1

ABSTRACT
Expansive soils shrink and swell when subjected to variations in moisture content, which often result fromseasonal climatic changes and the presence of nearby vegetation. These variations in moisture can create nonuniform ground movements that may lead to structural distress and increased roughness levels in road pavements.
Although the drying effect of roadside vegetation has always been expected to exacerbate moisture variation andassociated roughness levels in areas of expansive soil deposits, no studies have yet quantified this influence onroughness levels. In order to evaluate the relationship between roadside vegetation and pavement roughnesslevels, vegetation data was collected and analyzed against historical road roughness data for 109 sections (each
100 m in length) for a rural highway in Victoria, Australia. Compilation of the vegetation database consisted ofcollecting information such as; number and location of roadside trees, size of canopy area, tree height, and adensity rating for any grass present in the unsealed shoulder. The road roughness data set comprised of six
biannual sets of international road roughness indices. This paper describes the development of the roadsidevegetation database and includes procedures for extracting accurate information from satellite and street view
imagery. It also presents the relationship finding between the presence of vegetation and increased roughnesslevels for a rural highway pavement in Victoria, Australia, founded on expansive soils.
Keywords: Expansive Soils, Road Roughness, Roadside Vegetation, Pavement Deterioration

Gilgai formation was first discussed by Aitchison [8],but is best described in the Australian Handbook ofSoils [9]. In short, expansive soils are primarily
responsible for the formation of Gilgai terrain as it isthe deep shrinkage cracks that allow moisture topenetrate deep in the ground and promote deep
seated swelling, which initiates the formation of themounds.
In Victoria (Australia), approximately 50 per
cent of its land area is covered by moderately tohighly expansive soils [10], [11]. Moreover, aconsiderable amount of these soils have the potential
to develop Gilgai characteristics. This presents asignificant problem to the maintenance of the State’sroad network. Furthermore, as a significant amountof Victoria’s highway network incorporates the useof trees and vegetation into the road reserve toimprove the overall experience of the road user,
these roadside trees can exacerbate the developmentof Gilgai and associated roughness. Current
literature suggests that trees can influence groundmoisture conditions up to 10 m below the groundand horizontally up to 1.5 to 2 times the height of
the tree [11]. However, no studies have everquantified the relationship between the presence ofroadside trees and roughness levels.

INTRODUCTION
Expansive soils are common throughout
Australia with six out of the eight capital cities andsurrounding areas affected by such geologicalconditions [1], [2]. Expansive soils are defined as
those that shrink and swell when subjected tovariations in moisture. Factors that affect thisbehavior is the particle size distribution, type andquantity of clay present, and everything else that
controls ground moisture variation such as climate,site drainage (topography) and the presence ofvegetation. This shrink-swell action creates many
problems for light structures such as road pavements[3]. The presence of an expansive soil subgrade canincrease roughness levels (loss of surface shape) andlongitudinal cracks in the pavement structure.
Longitudinal cracking usually develops due to rapid(seasonal) desiccation and shrinkage of the soil
beneath the edge of the pavement [4], while roadroughness can be attributed to non-uniform volume
change beneath the pavement. Gilgai phenomenon isoften responsible for such non-uniform volume
changes [5]-[7].
Gilgai is an Aboriginal (Australian) word thatmeans “small waterhole”, which represents the waterthat collects in the depressions of undulating ground.
The general physical features of soils exhibiting
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GEOMATE- Bangkok, Nov. 14-16, 2016

Highway, consecutive road roughness data wasavailable from 1995 to 2009. From the 80 km of
Borung Highway selected, more than 150 sections(each 100 m in length) were identified as containing
high levels of roughness. These sections werelocated and referenced through the State Road
Referencing System (SRRS). This allowed
roughness progression rates (pre-maintenance andpost-maintenance) to be accurately calculated foreach 100 m section, of which 109 sections wereidentified as potential test sections for this study.
Pre-maintenance roughness progression rates refer tothe change in roughness (IRI) per year before
maintenance. Pre-maintenance sections were easily
identified through observation as any sudden drop in
IRI values clearly indicated maintenanceintervention.

SITE DESCRIPTION
To evaluate the effect of roadside vegetation onroad roughness levels, the Borung Highway inVictoria (Australia) was selected as: (i) it is locatedin a highly expansive (alluvial) geological area, (ii)
Gilgai formation is common, (iii) climatic conditions
promote significant shrink-swell behaviour, (iv) a
good distribution of varying roadside vegetation
exists, and (v) availability of quality historical roadroughness data. The section of Borung Highway
used in this study is located between the townshipsof Litchfield and Dimboola (approx. 80 kms inlength), which is shown in Fig. 1. In this figure, theyellow and orange zones represent the presence of
alluvial expansive soils, whereas the pink zones
indicate residual expansive soils. The degree of
expansive potential of the soil is indicated by thedepth of color. The Borung Highway is a two-lane
flexible rural highway that comprises of a thin
sprayed seal wearing course. In total, 109 sections(each 100 m in length) were evaluated in this study.

Tree Data Collection
In order to evaluate the relationship between
pavement roughness and the presence of roadsidevegetation, a roadside tree database was developed.
This database included information such as (i)number of trees on each side of the pavement, (ii)location of individual trees along the pavement
measured in SRRS chainages, (iii) offset distances
between the trees and centre-line of pavement, (iv)canopy area of each tree, (v) height of each tree, and(vi) density of grass present in the shoulder. To
achieve this, each tree was identified and was given
its own unique identification number. All of this data
was collected using satellite and street view imagery(Google Earth Pro). A schematic diagram showing
some of the variables measured for the roadside tree
database is presented in Fig. 2.

Fig. 1 Expansive soil map of Victoria (Australia)
vs location of test site [12]DATA COLLECTION AND METHODOLOGY
Data for this study was collected in two parts.
This included pavement roughness data and roadsidetree and vegetation data. The road roughness data
was provided by Vicroads (State Road Authority ofVictoria, Australia), whereas the roadside tree data
was mostly collected using satellite imagery.

Fig. 2 Elevation view of roadside trees withvariables measured
Number of trees

Pavement Roughness Data Collection
From satellite imagery (Google Earth Pro), thenumber of trees present in the road reserve caneasily be counted. Mature trees, young trees(including bushes) and even trees with no canopy
were all counted and input into the database. For
large groups of trees, street view imagery was also

Vicroads typically conduct detailed road surveys
once every two years across the State of Victoria,
which consists of measuring the longitudinal roadprofiles and calculating the International Roughness
Index (IRI) at 100 m intervals. For the Borung564

GEOMATE- Bangkok, Nov. 14-16, 2016used to help identify individual trees.
Offset distance between tree and pavement
The distance between each tree and centre-line ofthe pavement was measured using Google Earth Pro.
An example of this using the “Ruler” function isshown in Fig. 3. In this figure, the tree is identified
as DTLR104783 as it is located on left side at achainage of 104783 and has an offset distance of
17.6 m from the centre-line of the pavement.

Fig. 4 Example measurement of tree canopy area
using a polygon in Google Earth Pro
Tree height
To measure tree height, both satellite and street
view type images in Google Earth Pro need to be
used. First, satellite imagery is used to position asmall polygon near the base of the tree (using 3D
Polygon tab). A narrow or slender polygon ispreferable as it is easier to position (see Fig. 5a). The
sides of the polygon then need to be extended to theground using the scale, and the height of the polygon
adjusted. The style and color of the polygon can be
modified for better visibility depending on the image.
Street view is then used to adjust or fine-tune theheight (if required) of the polygon to match theheight of the tree (see Fig. 5b). The accuracy of thisapproach may depend on the initial positioning of
the polygon and visual adjustment of the height. The
closer the position of the polygon to the base of thetree, the greater the level of accuracy will beachieved. In Fig. 5, the height of the tree wasmeasured as 9 m.

Fig. 3 Example measurement of offset distance
between tree and pavement using Google Earth Pro
Canopy area
Similar to measuring distance, the canopy area
can be estimated in Google Earth Pro. The outline
and fill option should be used for accuracy. Propercare should also be taken while drawing the polygon
to not include any tree shadows, as this may result inover- estimating the area. Fig. 4 shows an example
of a polygon, which measures the tree canopy area
as 47.1 m2.

565

GEOMATE- Bangkok, Nov. 14-16, 2016of 5 referred to a shoulder fully occupied with grass.
It was decided to use this rating system toapproximate the density of vegetation (grass) in theshoulder and convert into a metric scale. The ratingof the shoulder vegetation was performed throughstreet view of Google Earth. An example of a street
view image with varying degrees of shoulder
vegetation density is shown in Fig. 6. Here, thedensity rating for grass present in each unsealed
shoulder was 2 and 1 for the left and right shoulders.
This is because grass (moderate) has started to
encroach into the left shoulder, while nearly no grass
was present in the right shoulder.

Fig. 6 Example density ratings of grass present ineach shoulder of the pavement using Google Earth
Pro

(a)

DATA ANALYSIS AND RESULTS
As the primary objective of this study was toevaluate and quantify the effect of roadside trees on
the development of road roughness, all measured
tree parameters were statistically evaluated as
independent variables and pavement roughness (interms of IRI) was set as the dependent variable.
Statistically, no correlation against the presenceof roadside trees (i.e. quantity, height, or location
with respect to the pavement) could be establishedagainst individual pavement roughness levels, whichwere based on single road surveys. This is most
likely due to the fact that road pavements naturally
deteriorate over time due to a number of different
factors, and the age of the pavement was notconsidered in this initial analysis. Therefore, toremove the age (or time factor) of the pavementfrom the analysis, we evaluated the roadside tree andvegetation variables against roughness progression
rates (RPR) instead.
The roughness progression rates were calculated
for pre-maintenance and post maintenance events,
and were based on change in IRI per year. Premaintenance roughness progression data consisted ofany change in roughness (per year) before any
noticeable effects from maintenance or rehabilitation.
While post maintenance roughness progression data(b)
Fig. 5 Example measurement of tree height using 3D
polygon tab in Google Earth Pro: (a) Satellite image,and (b) Street view image
Shoulder vegetation
To assess the effect of any grass present in theroad shoulder on road roughness, a rating system(using values between 0 and 5) was employed. A
zero rating indicated no grass present, while a value
566

GEOMATE- Bangkok, Nov. 14-16, 2016

consisted of any changes after a significant
maintenance event.
The results of the statistical analysis using
roughness progression rates (instead of individual
roughness levels) yielded much better correlations,which is presented in Table 1. The independent
variables evaluated were (i) Number of Trees (NoT),(ii) Average Distance between Trees and Pavement(AvDTrPv), (iii) Average Canopy Area (ACA), (iv)
Average Tree Height (ATH), and Average Grass
Density (AGD) rating. In this study, trees wereconsidered up to a AvDTrPv distance of 39.4 m, andthe average tree height ranged from 4 to 16.7 m. In
addition, the statistical analysis was performed inthree modes. First, only the roadside tree populationon the “near” side adjacent to the surveyed lane wasconsidered for the analysis. Then, only roadsidetrees on the “far” side were evaluated. Finally, allroadside trees (both sides) were taken into
consideration and evaluated with the lane roughness
progression rates.

This may result in greater moisture loss in themerged zone and create greater associated shrinkage
and roughness progression. This fact was supportedby the correlation results as when we considered theroadside trees on both sides of the pavement
together the significance of the correlations
improved yet again. The best correlation developedin this study was recorded between RPR and theaverage distance between tree and pavement(AvDTrPv) with a p value of 0.015. This was closely
followed by the correlation between RPR and theaverage tree height (ATH) with a p value of 0.02.
The Pearson coefficient of both of these correlations
was measured at 0.233 and 0.223 respective, whichindicated a small to medium effect [13].

Table 1 List of independent roadside vegetation
variables and associated correlations with overalllane roughness progression rates.
Independent
Variables
Near
Side
Trees

Far
Side
Trees

All
Trees

NoT
AvDTrPv
ACA
ATH
AGD
NoT
AvDTrPv
ACA
ATH
AGD
NoT
AvDTrPv
ACA
ATH
AGD

Significance
p
0.222
0.227
0.101
0.175
0.142
0.392
0.835
0.634
0.513
0.625
0.111
0.015
0.198
0.02
0.17

Pearson
r
0.118
0.117
0.158
0.131
0.141
0.083
0.02
0.046
0.063
0.047
0.154
0.233
0.124
0.223
0.132

Fig. 7. Definition and influence of nearside and far
side trees relevant to surveyed lane
Between RPR and AvDTrPv, a negative relationshipexists, which means the RPR decreases when the
AvDTrPv increases. Thus, the further the roadsidetree is from the pavement, the less effect it has onroad roughness. The equation for predicting RPRs
with respect to the average distance between theroadside trees and pavement is shown in Eq. (1).
𝑹𝑹𝑹𝑹𝑹𝑹𝑳𝑳𝑳𝑳𝑳𝑳𝑳𝑳 = 𝟎𝟎. 𝟐𝟐𝟐𝟐𝟐𝟐 − 𝟎𝟎. 𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎(1)𝑹𝑹𝑹𝑹𝑹𝑹𝑳𝑳𝑳𝑳𝑳𝑳𝑳𝑳 = 𝟎𝟎. 𝟐𝟐𝟐𝟐𝟐𝟐 − 𝟎𝟎. 𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎(2)

Between RPR and ATH, a negative relationshipwas also found to exist. This suggests that thegreater the height of the tree, the lower theroughness progression. This finding seems unusual
at first as it is commonly believed that the bigger thetree, the more widespread the root system, and thus
the greater the influence of moisture variation.
However, this finding implies that the small the tree,
the greater the influence. This could be due tosmaller trees being younger trees, which havegreater growth rates and thus require more moisture.
The equation for predicting RPRs with respect to theaverage tree height is shown in Eq. (2).

Results showed that when we considered trees on
only one side of the pavement; the correlation was
not significant for any of the independent variables
as all p values exceeded 0.05. The correlation results
when considering the nearside trees only (although
not significant) was much better than the correlation
results when only the far side trees were considered.
The reason for this is simply due to the near sidetrees being much closer to the surveyed pavement
lane. Fig. 7 defines the nearside and far side trees
with respect to the surveyed lane. From the results inTable 1, it is clear that the further the tree is from thepavement, the less influence it has on the RPR.
Furthermore, when trees are located on both sides ofthe pavement the drying influence zones may merge.
567

GEOMATE- Bangkok, Nov. 14-16, 2016

[2] Fityus S, Smith D, and Allman M, “Expansive

When both of these independent variables were
evaluated together, the significance improved
marginally as the Pearson coefficient increased to0.302 and thus became a medium effect on the RPR.
The equation for predicting RPRs with respect to a
combination of AvDTrPv and ATH is shown in Eq.
(3).

soil test site near Newcastle,” Journal of
GeotechnicalandGeoenvironmental
Engineering, Vol. 130, 2004, pp. 686-695.
[3] Krohn

JP,

SlossonJE,“Assessment

expansive soils in the United
Expansive Soils, 1980, 99.596-608.

𝑹𝑹𝑹𝑹𝑹𝑹𝑳𝑳𝑳𝑳𝑳𝑳𝑳𝑳 = 𝟎𝟎. 𝟑𝟑𝟑𝟑𝟑𝟑 − 𝟎𝟎. 𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎 − 𝟎𝟎. 𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎𝟎 (3)ofStates,”[4] Luo R, and Prozzi JA, “Development of
longitudinal cracks on pavement over shrinking
expansive subgrade,” Road Materials andPavement Design, Vol. 11, 2010, pp. 807-832.

CONCLUSIONSThis study revealed that the presence of roadsidetrees do have some salient effect on the developmentof roughness for flexible pavements located in areasof alluvial expansive soils. Although no relationshipbetween the studied tree parameters and individual
roughness levels could be established, weak to
medium correlations were verified when based onroughness progression rates. The best correlations
were found between roughness progression rates and(i) the relative distance between the tree andpavement, and (ii) the height of the tree.
Surprisingly, no correlation was able to beestablished between roughness progression and thequantity of trees present in the road reserve or the
canopy area.
As road roughness was measured using theInternational Roughness Index (IRI) in this paper, itis planned to further examine the effect of these
physical tree parameters on other roughness indices,such as those derived through the Butterworth Filter
or Power Spectral Density (PSD). By evaluating
roughness progression within the individual
wavebands, it is anticipated that a better prediction
model to quantify the effect of roadside vegetation
can be determined.

[5]

Evans RP, Arulrajah A, and Horpibulsuk S,“Detecting gilgai relief beneath sealed flexible
pavements using road profile and roadroughness measurements,” Indian Geotechnical
Journal, Vol. 45, 2015, pp. 431-440.

[6] Lytton RL, Boggess RL, and Spotts J,“Characteristics of expansive clay roughness ofpavements,” Transportation Research Record,
1976.
[7] Velasco MO, and Lytton RL, “Pavement
roughness on expansive clays,” Transportation
Research Record, 1981.
[8] Aitchison G. “The mechanics of gilgai
formation”, in Proc. of Conference on Soil
Science, 1953, pp. 1-3.
[9] Hallsworth E, “A handbook of Australian soils,”
Adelaide, Rellim Publications, 1968.
[10] Holland J, and Richards J, “Road pavement on
expansive clays,” Australian Road Research,
Vol. 12, 1982.
[11] Richards B, Peter P, and Emerson W, “TheACKNOWLEDGEMENTS

effect of vegetation on the swelling andshrinking of soils in Australia,” Geotechnique,Vol. 33, 1983, pp. 127-139.

The authors acknowledge the support offered by
Vicroads in providing the raw longitudinal roadprofile measurements, which was used to calculate
the roughness indices.

[12] Evans R, “Synchronisation of road profile data
and the assessment of road roughness using
waveband analysis,” PhD Thesis, Department
of Civil and Construction Engineering,
Swinburne
UniversityofTechnology,
Melbourne, Australia, 2013.

REFERENCES[1] Delaney M, Li J, and Fityus S, “Field
monitoring of expansive soil behavior in the
Newcastle-hunterregion,”JournalofAustralian Geomechanics, Vol. 40, 2005, pp.
3-14.

[13] Field A, “Discovering statistics using SPSS,”
Sage Publications, 2009.

568

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

INFLUENCE OF WEATHERING OF BOTTOM ASH ON THE
LEACHING BEHAVIOR OF CESIUM
Yasumasa Tojo1, Saori Iwamoto1, Mikako Ishii1, Toshihiko Matsuto1 and Takayuki Matsuo1
1
Graduate school of Engineering, Hokkaido University, Japan

ABSTRACT
After nuclear accident in 2011, incineration residue contaminated by radioactive cesium had been produced.
Government decided in 2011 that incineration residue with less than 8000 Bq/kg of radioactivity could be
disposed of into conventional MSW landfill. Since radioactivity of bottom ash was reported to be low comparedwith fly ash, it had been disposed of MSW landfill based on the government regulation. However, long-termleaching behavior of cesium contained in bottom ash when being disposed of into landfill was not known. Thus,in this study, phase transformation and leaching of cesium from incineration bottom ash were investigated byconducting accelerating weathering experiment. Four kinds of weathering condition were set; namely, blank(under nitrogen gas condition), wetting and drying, freezing and thawing, and exposure to carbon dioxide gas.
Leaching of cesium under blank condition didn't change. On the other hand, drastic decrease of leaching of
cesium was identified under carbon dioxide exposure. From XRD analysis, formation of calcite was identified. It
was deemed that newly formed calcite inhibited the leaching of Cs by covering reactive surface of bottom ash. In
both wetting-drying and freezing-thawing condition, leaching of Cs increased initially. But it decreased after
several weeks then it became lower than blank finally. This decrease of leaching also seems to be caused byformation of calcite. Besides, in order to confirm the assumption in which the restrain of cesium leachingoccurred by calcite formation, SEM-EDS analysis was performed. By the analysis, calcite formation on thesurface was clearly identified.
Keywords: Incineration Bottom Ash, Cesium, Leaching, Carbonation
Therefore, it is not certain that Cs keeps being stably
captured in the ash for long-term after disposal into
MSW landfill. Thus, in this study, the effect ofweathering of the bottom ash on the leaching
behavior of Cs was investigated in order to confirm
their long-term stability in landfill.

INTRODUCTION
After the accident of Fukushima-Daiichi nuclear
power plant in 2011, municipal solid waste
incinerators in east Japan generated ashes containing
radioactive cesium (Cs) in high concentration.
Government decided that the waste of 8000 Bq/kg orless could be disposed of into MSW landfill as
before [1]. Radioactivity of fly ash was very highand special measure is necessary for its disposal. On
the other hand, radioactivity of bottom ashes was
low and many of them satisfied the government’s
criteria so that its disposal to MSW landfill has been
carried out. National Institute for Environmental
Studies was conducted sequential extraction test forvarious kinds of waste that were contaminated byradioactive cesium. And they reported that watersoluble form of cesium in bottom ash is only about
3% and approximately 80% of cesium existed as
residual form (i.e. the last fraction of extraction
tests) [2]. Their results indicated that possibility ofcesium leaching was low even if the bottom ash wasdisposed of into ordinary MSW landfill sites.
However, it is well known that the incineration
residue is unstable substance and its mineral phases
are changed by various environmental factors.
Especially, leaching behavior of inorganic elementschanges by weathering along with elapsed time.

MATERIALS AND METHODS
Materials
In general, the total cesium content (stable
cesium and radioactive cesium) in the incineration
bottom ashes discharged from a real incinerator isvery small, and moreover the content of radioactive
cesium is also very small. For instance, total cesium
content, which is comprised of mainly stable-cesium,
is reported to be less than several mg/kg [3]. As forthe content of radioactive cesium, it is estimated to
be several ng/kg, if radioactivity of the incineration
ash is assumed to be several thousands Bq/kg. These
values imply that chemical analysis of cesium
contained in the incineration residue is quite difficult,especially radioactive cesium. So, in this study,
artificially created ash, of which stable cesium
content was made quite high, was used. The ash was
created by adding cesium carbonate (Cs2CO3) toRDF (Refuse-Derived Fuel) and combusting it by569

GEOMATE- Bangkok, Nov. 14-16, 2016

pilot-scale stoker incinerator of which capacity was
3 tons/day [4]. After combustion, the bottom ash
was cooled under ambient temperature. Since, thefurnace allows collecting fly ash and bottom ash
separately, the bottom ash doesn’t contain fly ash.
After creation, 20 kg of the bottom ash was taken
to the laboratory. Immediately, the ash was
pulverized to less than 0.5 mm by jaw crusher(Retsch, BB-50) to obtain uniform sample. At first,
the metal content was measured by aqua-regia
digestion and the atomic absorption spectrometry inorder to elucidate characteristics of this artificiallycreated bottom ash. Moreover, to determine theamount of soluble cesium, the serial batch leaching
test was executed [5]. In the serial batch test,
leaching test in L/S (liquid to solid ratio) =10 was
repeated five times. Fig. 1 shows the content of Cs,
K, Na, and Ca.
Hardry soluble fraction

Percentages of each fraction

100%
1770

3335

Methods of Weathering Acceleration
The Four different conditions (Blank, Drying andwetting, Freezing-thawing, and CO2 gas exposure)
were established for the weathering experiment.
Blank: 10g of pretreated ash samples was put
into the 50mL screw bottle. To prevent any reaction
and so as not to cause weathering, N2 gas waspurged into the head space. After capping, it was left
sitting at the room temperature. To prevent any
reaction as much as possible, the N2 purge was done
once a week. This system was created for thecomparison with another three series.
Drying and wetting: 10g of the pretreated sample
was filled in the small crucible and then the crucible
was fixed in a steel tray that spread the glass bead.
Distilled water was added to the sample to become
saturated water content and it was left sitting for 30
minutes. Then, it was heated in the dryer for 1.5
hours at 100˚C. This process was repeated threetimes a day and it was continued for eight weeks.
This series was made in order to accelerate physicalweathering (i.e. physical collapse of the particle
structure of the ash) by repetition of drying andwetting.
Freezing and thawing: 10g of the pretreatedsample was filled in the small crucible. After adding
distilled water to the sample to make its saturated
water content, the crucible was capped and it wascooled in the freezer for 1.5 hours. After that, thefrozen sample was heated with the crucible for one
hour at 40˚C in the dryer, and melted. This process
was repeated three times a day and it was continued
for eight weeks. This series was made in order toaccelerate physical weathering (i.e. physical collapseof the particle structure of the ash) according to the
expansion and shrinkage of pore water by thefreezing and melting.
The CO2 exposure: 10g of pretreated sample was
spread in the steel tray and then distilled water wasadded to make its water content at field capacity.
The tray with sample was put into the air-tightdesiccator and CO2 gas was filled in it. To keep thehumidity in the desiccator, steel tray filled with
distilled water was placed inside. The moisture
content was monitored everyday and readjusted to
the field capacity. Besides, CO2 gas was always kept
injected into the inside. This series was made inorder to accelerate chemical weathering (i.e.
chemical transformation by contacting with CO2).
This is because the major weathering reaction
occurring in the incineration ash is reported to be thecarbonation [6].

Soluble fraction
2776

21695

37268

284392

80%
60%
40%

6631

21097

20%
0%

Fig. 1

Content of hardly soluble and soluble
fraction on Cs, K, Na, and Ca in artificiallycreated bottom ash. Number in the barshows each content (unit: mg/kg-ash).

In the figure, content of each element is indicatedby distinguishing the soluble fraction and hardly
soluble fraction. Although metal content and theamount of soluble fraction were almost the samewith that of general incineration ashes, soluble
fraction of Cs was 20% and it was higher than the
one that had been detected in actual furnaces.
Because hardly soluble Cs occupied the majority of
the content in the ash actually generated from theincinerator in East Japan, existence of soluble Cs is
undesirable in the examination of a long-term
behavior of Cs. Therefore, the ash used for theweathering acceleration experiment was washed in
L/S=10 beforehand. Amount of soluble Cs becomesto about 1% in total content after the pretreatment.
And it was not below the detection limit. Thus, it
was regarded as suitable to pursue the change in theleaching amount.

570

GEOMATE- Bangkok, Nov. 14-16, 2016

Eight pieces of sample were made for one
weathering series. The weathering acceleration was
done at the same time by above-mentioned each four
methods. One piece was collected every week, andthe sample was subjected to the analysis.

minerals contained in the ash. Before analysis,
samples were dried at 80˚C and then they weremilled. X-ray diffractometer (Rint-2000, Rigaku)with Cu-Kα radiation was used. Scan range was 580˚ with scan speed 4˚ per minute.

Leaching Test Procedure

Microscopic Analysis of Weathered Ash

2g of weathered sample was put into Erlenmeyer
flask and 20mL of distilled water was added(L/S=10). The flask was shaken for six hours byusing the desk-top shaker. It was reported that there
are two types of Cs leaching from soil (i.e. Cs
absorbed with soil colloid and free ion state). Thus,in this study, the leaching liquids were filteredthrough two stages. At first, it was filtered by 1µm
glass-fiber filter and then a part of the filtrate wasused for the analysis. After that, remainder filtrate
was filtered by 0.1µm membrane. Both liquids wereanalyzed by frame atomic absorption spectrometry(AAS: Hitachi Z-8200). Analyzed elements were Cs,
K, Na, and Ca.

In the weathering experiment, obvious inhibitionof Cs leaching was confirmed in the sample of CO2
exposure condition. Thus, by using weathered
sample of the CO2 exposure and blank, thin sections
were prepared. Thin sections were observed by
Scanning Electron Microscope - Energy Dispersive
X-ray Spectroscope (SEM-EDS: JEOL JSM6360LA) and elemental distribution was investigated.
Besides, to analyze formation of newly formed
minerals in weathered ash samples, Polarizing
microscopes （OLYMPUS CX31-P）was used.
RESULTS AND DISCUSSION
Change of Leaching Behavior of Cs by
Weathering

Identification of Minerals

The leaching amount of Cs from the ash exposed
to each weathering condition is shown in Fig. 2.
Though the direct result obtained by AAS is theconcentration of the leaching liquid, the leaching

The Samples after two weeks of weathering were
subjected to the Powder X-ray Diffraction (XRD)
analysis in order to investigate changes of major
Amount of leaching [mg/kg]

300

300

Blank (N2)

1µm filtration250

Wet-Dry250

0.1µm filtration

200

200

150

150

100

100

50

50

0

0
0

10

20

30

40

50

60

0

10

20

Elapsed days
300

Amount of leaching [mg/kg]

30

40

50

60

Elapsed days
300

Freeze-thaw

250

Carbonation (CO2)

250

200

200

150

150

100

100

50

50

0

0
0

10

20

30

40

50

60

Elapsed days

0

10

20

30

40

50

Elapsed days

Fig. 2 The amount of Cs that leached out from the weathered bottom ash at each elapsed day
571

60

GEOMATE- Bangkok, Nov. 14-16, 2016

amount, which is converted from the concentrationby using L/S ratio, is indicated. Each panel shows
the result of the sample that was exposed to eachweathering condition. In addition, the results of two
filtration stages are shown by separate line. As
indicated in the figure, the leaching amount obtainedby both filtration stages didn't differ much.
Concentration of the filtrate obtained by 1um filter isslightly high. This implies that Cs leached out in thisstudy was mainly ionic form.
Leaching amount of Cs from blank sample (i.e.
N2 exposure) didn't change and it kept almost
constant value similar to the initial for through entire
experimental period. On the other hand, leaching ofCs from sample exposed to CO2 circumstancedrastically decreased within a week as indicated in
the lower right panel. It reached to below thedetection limit at two weeks. This rapid decrease isthought to be caused by the formation of calcite at
the surface and it may inhibit the Cs leaching bycovering the ash particle. The result obtained fromboth drying-wetting and freezing-thawing conditions
shows slight increase of Cs leaching in the first week.
The increase of leaching amount was almost 2.5
times to the result of blank. This increase is thought
to be caused by collapse of the ash particle structure,
which captures Cs to the inside of it, by the physicalaction due to the thermic effect or the
expansion/contraction of pore space. For example,
Saffarzadeh [7] reported that Cs is captured inglassy-amorphous phase or some alminosilicates.
Thus, if these structures are affected by weathering,leaching of Cs seems to be possible. However,subsequently, the leaching of Cs from bothweathering conditions showed the decreasing trend.
Especially, Cs leaching from the ash subjected to thedrying-wetting condition became lower than it of theblank. This means that the soluble Cs contained in
the ash was transformed to the hardly soluble Cs.
The conceivable reason of this trend is the
carbonation similar to the CO2 exposure condition.
In both weathering conditions, the ash is exposed tothe atmosphere. As for the wetting-drying condition,
the ash is exposed to the atmosphere during drying.
In the freezing-thawing, the situation is the same.

to be occurred in the weathered sample and calcite
formed. This calcite formation might affect theleaching of Cs.

Carbonation (CO2)Freeze-thaw

Wet-Dry
Blank (N2)0Fig. 32040

2θ [° ]

60

80

Powder XRD diffractogram on the eachweathered bottom ash samples (after two
weeks). Rectangle drawn by dashed line
encloses main peak of calcite.

Table 1 Change of main peak height of minerals
identified each bottom ash samples.
MineralBlank

Gehlenite
QuartzLime
CalciteHematite

1327
1381
193
438
235

Wetdry
1575
1198
206
851
201

Freezethaw
1290
579
181
784
185

CO2
Exposure
1503
965
196
1903
121

Microscopic Analysis on the mechanism of theleaching inhibition of Cs
Since calcite formation in weathered ash could
be confirmed from the result of XRD, microscopic
analyses were performed to verify the existence of
calcite and its location. At first, some ash particles
were observed at a low magnification by using
polarizing microscope. Images observed by thetransmissive light mode on the blank and the CO2
exposed sample are shown in Fig. 4. Some ash
particles exposed to CO2 can be seen in panel (b). It
seems that the perimeter of the particles shines inyellow. When the observed object emits various
colors depending on the rotation angle in thepolarizing microscope analysis, the object isconsidered to be some crystal. This is because
mineral in general is composed of some crystal
structure and it has polarization property. Therefore,
the image of these particles shows that some sort ofmineral formed in the sample exposed to CO2.

Change of Mineral Phases by Weathering
Minerals identified in XRD analyses which wereconducted for each sample after two weeks areindicated in Table 1. The values in the table are thepeak height of these minerals. Moreover, the X-ray
diffraction charts of each sample are also shown in
Fig. 3. In the figure, main peak of calcite is enclosedby the rectangle of dashed line. The peak height ofcalcite in weathered sample except for the blankincreased in common. Especially, the height of theCO2 exposed sample became five times higher thanthe blank. From these results, carbonation is deemed572

GEOMATE- Bangkok, Nov. 14-16, 2016(a)

peripheral region. This implies that some sort ofcalcium mineral surrounded the particle comprisedof silicon.

(b)Particle-1

Fig. 4

Observed image of weathered ash particles
by polarizing microscope; (a) blank, (b)CO2
exposure.

Next, a particle (Partilce-1) shown in Fig. 4 (b)(enclosed by rectangle) was focused and observedwith much higher magnification. As indicated in Fig.
5, obvious luminescence at the perimeter of theparticle can be confirmed.

Fig. 6

Magnified image (backscattered electron
image) of a region-1 in Fig. 4 and EDS
analyzing line and points.

Fig. 7

EDS mapping image (red: calcium, green
silicon)Region-1

Fig. 5Magnified image of a particle-1 in Fig. 4.

In order to identify the kind of the substanceformed at the perimeter of the particle, a region(Region-1) which is enclosed by rectangle in Fig. 5
was observed by SEM. Figure 6 shows thebackscattered electron image of the Region-1. Theedge of the particle is located at slightly right side of
the center. In this image, existence of some sort ofsubstance at the edge of the particle is obvious and it
is seen as pale gray color. This substance seems to
be formed as if it surrounds the particle.
Moreover, EDS analysis was performed in thisregion in order to know the element distribution.
Element mapping image of calcium and silicon isindicated in Fig. 7. Much Ca is distributed at the
peripheral region of the particle. The location where
Ca is abundant coincides with the gray color part in
Fig. 6. Furthermore, this region corresponds to theperimeter which shines in yellow in the image ofpolarizing microscope. On the other hand, as for the
distribution of Si, its intensity is high at the inside of
the particle. And Si is not much distributed at the

The result of EDS line analysis on line001 in Fig.
6 is shown in Fig. 8. Peak intensities of calcium andsilicon along with the distance are indicated. Peakintensity of Si is higher than Ca until 2mm ofdistance from left side. However, when the distanceexceeds 2mm, their intensities become reverse. The
point of reversal corresponds to the gray color partin the BEI (Fig. 6). And also it corresponds to thearea where the Ca is abundant in the element
mapping image (Fig. 7). Results of EDS point
analysis are shown in Table. 2. When focusing on
Ca, it was 9.5% at the point 004 that is the inside of
the particle. However, at the point 002 that locates atthe perimeter, percentage of Ca increased to 26.5%.
From all of the results mentioned above, it can beconcluded that some sort of calcium mineral formedat the perimeter of the ash particle which is exposedto the CO2 as the weathering acceleration.

573

GEOMATE- Bangkok, Nov. 14-16, 2016leaching of Cs by covering surface of bottom ash.
Incineration bottom ash is highly alkaline materialand carbonation promptly occurs after disposal.
Since the pH of the ash landfill is reported to be kept
high for long-term, Cs in bottom ash will be kept
stably by calcite for a long term.

300

Si

Peak intensity [-]

250
200
150

Ca

100

ACKNOWLEDGEMENTS

50

This research was supported by an Environment
Research and Technology Development Fund of the
Ministry of the Environment, Japan. Besides, manyof microscopic analyses conducted in this work wereassisted by Hokkaido University Microstructure
Characterization Platform’s facility allocated in
Nanotechnology Platform Program of the Ministryof Education, Culture, Sports, Science andTechnology (MEXT), Japan.

0
0

0.05

0.1

0.15

0.2

0.25

0.3

0.35

Distance [mm]Fig. 8Result of EDS line analysis on line001 in
Fig. 6.

Table 2 Result of elemental analysis by Energy
Dispersive X-ray Spectroscopy (EDS) on
points indicated in Fig. 6.

REFERENCES

Point
002
003
004
C
23.7
19.9
19.4
O
42.5
48.9
36.1
Al
1.1
1.1
8.2
2.3
2.7
15.8
Si
Ca
26.5
25.2
9.5
Other
3.9
2.3
11.0
Total
100
100
100
Numbers in the table indicate percentage of eachelement (%).

[1] Japan Ministry of Environment, “Managementof off-site Waste Contaminated with
Radioactive Materials due to the Accident at
Fukushima Nuclear Power Stations”, 2012,
http://www.env.go.jp/en/focus/docs/files/20121
128-58.pdf.
[2] Japan National Institute for Environmental
Studies, “Proper treatment of wastes
contaminated by radioactive substances, 2012,
http://www.nies.go.jp/shinsai/techrepo_publicve
r_e_121128.pdf, 2012.
[3] Landsberger S, Buchholz B. A, Kaminski M,
Plewa M, “Trace elements in municipal solid
waste incinerator fly ash”, Journal ofRadioanalytical and Nuclear Chemistry, Vol.
167(2), 1993. pp 331-340.
[4] Iguchi T, Kawano T, Fujihira H, “Combustion
Behavior of Cesium in MSW Incineration
Process”, Proceedings of the 1st IWWG-ARB
Symposium, 18-21 March, 2013.
[5] Sakanakura H, Osako M, Kida A, “Effect of
exposure test conditions on leaching behavior of
inorganic contaminants from recycled materials
for roadbeds”, Waste Management, Vol.29 (5),2009, pp.1658-1665.
[6] Meima J.A, Comans R.N.J, “The leaching oftrace elements from municipal solid waste
incinerator bottom ash at different stages ofweathering”, Applied Geochemistry, Vol.14,
1999, pp.159-171.
[7] Saffarzadeh A, Shimaoka T, Kakuta Y, Kawano
T, "Cesium distribution and phases in proxyexperiments on the incineration of radioactively
contaminated waste from the Fukushima area,
Journal of Environmental Radioactivity,
Vol.136, 2014, pp.76-84.

It is difficult to determine concretely that theformed mineral is calcite from the element ratio
identified by EDS. However, as mentioned earlier,the peak of calcite was increased in XRD analysis.
Besides, the trend of Cs leaching inhibition is mostremarkable in the ash subjected to CO2 exposure.
Based on these facts, the possibility of calcite
formation at the surface of the ash particle is quite
high. And it will work to restrict Cs leaching fromthe incineration ash disposed of into MSW landfill.
CONCLUSION
In this study, to confirm long-term stability of
Cs contained in incineration ash disposed of into
MSW landfill, leaching of Cs from ash and phasetransformation were investigated by conducting
accelerating weathering experiment. Drastic
decrease of the Cs leaching was identified undercarbon dioxide exposure. From XRD analysis,formation of calcite was identified. By microscopic
analysis by using SEM-EDS and polarizing
microscope, newly formed mineral that is comprisedof Ca mainly was identified at the perimeter (i.e.
surface) of the ash particle. Thus, it deemed that
newly formed mineral is calcite and it inhibited the574Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THERMOELECTRIC POWER (TEP) MEASURMENT OF GELGROWN BARIUM OXALATE CRYSTAL
Paresh Vasantlal Dalal1
1

Physics Research Lab, Shri V. S. Naik Arts, Commerce and Science College, Raver-425508, India

ABSTRACT: An attempt is made in the present work to characterize gel grown barium oxalate crystals by
Thermo Electric Power (TEP) measurement. Different parameters such as Fermi energy and mode of
scattering were calculated. To calculate Fermi energy and scattering parameter of a material, a graph ofSeebeck coefficient(S), versus reciprocal of temperature difference (1/ΔT) is plotted. The slope of the graphis -27.50 mV and intercept is 0.181 mV/K, and hence Fermi energy, E F = 0.028 eV. Scattering parameterhas calculated 0.4. The experimental value obtained for A = 2.10 is in well agreement to conclude that the
conduction of heat in the material may be due to the lattice or phonons and can be associated with lattice orphonon scattering.

Keywords: Barium oxalate, Thermo Electric Power (TEP), Fermi energy, Mode of scattering(Maharashtra). A single crystal of barium oxalate
was placed in the instrument at a proper position
and made all equipment ready for working. One
end of the crystal was heated with the heating
device upto certain fixed temperature. The
temperature difference occurred between two ends
and corresponding developed e.m.f. were recorded.

1. INTRODUCTION
Thermoelectric power (TEP) has attracted
many researchers for its applications in designing a
portable refrigerator [1], and power generations.
Narrow gap Bi 2 Te 3 semiconductors and its alloys,
are found to be the most efficient thermo-electricmaterials [2]–[5], as its Figure of merits (Z =S2/kρ, where S is Seebeck coefficient, k is thermal
conductivity and ρ is resistivity) have highSeebeck coefficient and therefore they have highdemand in electronics field. Many insulators haveshown the characteristics of exhibiting large TEP.
Insulators have extremely high resistance;
therefore conduction can take place by several
mechanisms such as thermal or Schottky emission,tunneling, Poole-Frenkel effect, field emission,
space charge limited conduction etc.[6].
Measurement of thermoelectric power has
distinctive advantages over other methods because
measured thermoelectric voltage is directly related
to the carrier concentration [7]. Many researchers
have reported thermoelectric power measurementof semiconducting materials [8]-[9], and some onorganic materials [10]-[17]. However, there are nosuch studies are reported on oxalate material.
Looking at the importance of thethermoelectric power (TEP), author of this paper
has therefore carried out a systematic investigationof gel grown barium oxalate crystals. Different
parameters such as Fermi energy and mode of
scattering were also calculated.

3. OBSERVATIONS AND RESULTS
Employing electrical heating device to one endof a crystal produces a temperature gradient, (ΔT)
between two ends of a material and give rise to an
e.m.f. This developed e.m.f. is known as thermo
e.m.f (ΔE), which is directly proportional to ΔT.
The temperature difference, ΔT between two ends
of a crystal is ranging within a limit of 9 to
12K.Thus, Seebeck coefficients ‘S’ at different
temperatures were calculated by using theequation:
S= (ΔE/ΔT)
The variation of thermo electric power, orSeebeck coefficient ‘S’ with respect to absolutetemperature is shown in Fig. 1.
To calculate Fermi energy and scatteringparameter of a material, a graph of Seebeckcoefficient(S), versus reciprocal of temperature
difference (1/ΔT) is plotted and shown in Fig. 2.
The slope of the graph is -27.50 mV
And intercept is 0.181 mV/K
So the Fermi energy,=(slope X charge on an electron)EF=
27.50 x1.6x10-19 joules=
27.50 meV=
0.028 eV
(intercept)
Constant A =meV/K

2. EXPERIMENTAL
Thermo electric power of barium oxalate single
crystal was measured on an instrument assembled
in the Physics Research Lab, Shri V. S. Naik Arts,
Commerce and Science College, Raver.

𝐾𝐾𝐵𝐵

Where K B is Boltzmann constant575

GEOMATE- Bangkok, Nov. 14-16, 2016

Constant A =(0.181X10−3)eV/K(8.617X10−5) eV/K

phonons and can be associated with lattice orphonon scattering [6] (Table 1).

Seebeck coefficient 'S' in mV/ K

A
=
2.10
Scattering parameterαs =(5/2) – 2.10=
0.4

Table 1 Values of A and their corresponding
Mode of scattering

0

Scattering mode
Ionized impurity
Piezoelectric (ioniclattice)
Grain boundary
Lattice or phonons
Vibrations at
Constant frequency
Ionic latticeDegenerate system

-0.5
-1

-1.5
-2

-2.5
-3

-3.5
-4

-4.5

Constant A
4
3
2
1
0.5
2.5
-

-5

300

320

340

360

380

400

4. CONCLUSION

Temperature T in Kelvin

In the present study, gel grown barium oxalate

Fig. 1: Plot of Seebeck coefficient (S) versus
Temperature(T)single crystal was characterized by Thermo electric

Seebeck coefficient (S) in mV/K

power (TEP) measurement. From the above
studies, following points are observed:-3
Y = -27.50x + 0.181-4

1. Calculated Fermi energy E F = 0.028 eV-5

2. Calculated Scattering parameter α s = 0.4-63. It is concluded that the conduction of heat in-7

the material may be due to lattice or phonon-8

4. And, can be associated with lattice or phonon
0

0.1

0.2

0.3

0.4

scattering.

1/∆T in (/K)
Fig.2:

5. ACKNOWLEDGEMENTS

Plot of Seebeck coefficient ‘S’ versus
Reciprocal of Temperature

The author is thankful to the University grants
Commission, New Delhi for providing financial
assistance under Minor Research Project Scheme.

3.1 Discussion
The value of Fermi energy E F is obtained fromthe slope of a graph and the constant A obtained

6. REFERENCES

from the intercept on Y-axis. The value of constant[1] Mahan G. D., “Good Thermo electrics”, Solid
State Physics, Vol. 51, 1998, pp. 81-157.
[2] Heikes R. R., and Ure R. W.,“Thermoelectricity: Science and Engineering”(Interscience,New York), 1961, pp. 576.
[3] Rosi F. D., “Thermoelectricity andthermoelectric power generation”, Solid-StateElectron., Vol.11, 1968, pp. 833-848.

‘A’ gives valuable information regarding the modeof scattering occurred in the material during the
conduction of heat.
The experimental value obtained for A, 2.10 isin well agreement to conclude that the conductionof heat in the material may be due to the lattice or

576

GEOMATE- Bangkok, Nov. 14-16, 2016

[4] Wood C., “Materials for thermoelectric energy
conversion”, Rep. Prog. Phys., Vol. 51, 1988,
pp.459-539.
[5] Goldsmid H. J., and Douglas R. W., “The useofsemiconductors
inthermoelectric
refrigeration”, Brit. J. Appl. Phys., Vol. 5,
1954, pp. 386.
[6] Goswami A., “Thin film Fundamentals”, New
Delhi: New Age International (P) Limited,
1996, pp.314.
[7] Brixner L. H., “X-ray study andthermoelectric properties of the W x Ta 1-x Se 2 ”,
J. Electrochem Soc, Vol.110, 1963, pp. 289.
[8] Calliat T., Fileurial J. P., and Borshchevsky A.,“Preparation and thermoelectric properties ofsemiconducting Zn 4 Sb 3 ”, J. Phys Chem Solid,Vol. 58(7), 1997, pp. 1119.
[9] Static V., Pierre A. C., Etsall T. H., and Mikula
R. J., “Preparation of tungsten sulfides by solgel processing”, J Non-Cryst Solids, Vo. 220,
1997, pp. 58.
[10] Wei, Q.S.; Mukaida, M.; Kirihara, K.; Naitoh,Y.; Ishida, T., “Thermoelectric power
enhancement of PEDOT:PSS in high-humidity
conditions”. Appl. Phys. Express, Vol. 7, 2014,
031601.
[11] Wei, Q.; Mukaida, M.; Kirihara, K.; Ishida, T.,“Experimental studies on the anisotropic
thermoelectric properties of conducting
polymer films”, ACS Macro Lett., Vol. 3, 2014,
pp. 948–952.

[12] Chang, K.-C.; Jeng, M.-S.; Yang, C.-C.; Chou,
Y.-W.; Wu, S.-K.; Thomas, M.; Peng, Y.-C.,“The thermoelectric performance of poly(3,4ethylenedioxythiophene)/poly(4styrenesulfonate) thin films”, J. Electron. Mater., Vol. 38,
2009, pp. 1182–1188.
[13] Zhang, B.; Sun, J.; Katz, H.E.; Fang, F.; Opila,
R.L. “Promising thermoelectric properties ofcommercial PEDOT:PSS materials and their
Bi2Te3 powder composites”, ACS Appl.
Mater. Interfaces, Vol. 2, 2010, pp. 3170–
3178.
[14] Kim, G. H.; Shao, L.; Zhang, K.; Pipe,
K.P.,“Engineered
dopingoforganic
semiconductors for enhanced thermoelectric
efficiency”, Nat. Mater. Vol. 12, 2013, pp.
719–723.
[15] Park H., Lee S. H., Kim F. S., Choi H. H.,
Cheong I. W., “Enhanced thermoelectric
properties of PEDOT:PSS nano films by achemical dedoping process”, J. Mater. Chem.
A, Vol.2, 2014, pp. 6532-6539.
[16] McGrail, B.T.; Sehirlioglu, A.; Pentzer, E.
“Polymer composites for thermoelectric
applications”, Angew. Chem. Int. Ed., Vol. 54,
2015, pp. 1710–1723.
[17] Yoshida, A.; Toshima, N., “Gold nanoparticle
and gold nanorod embedded PEDOT:PSS thinfilms as organic thermoelectric materials”, J.
Electron. Mater. Vol. 43, 2014, pp. 1492–
1497.

577

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

IMPACT ON AIR QUALITY BY INCREASE IN AIR POLLUTANTEMISSIONS FROM THERMAL POWER PLANTS
Akira Kondo1, Hikari Shimadera1 and Mai Chinzaka1
1

Osaka University, Japan

ABSTRACT: The amount of thermal power generation has increased significantly in Japan since the Great
East Japan Earthquake in 2011, resulting in increase in emissions of air pollutants. This research evaluatedthe impact of the emission increase on air quality in Kinki region, Japan by using the Community Multiscale
Air Quality model (CMAQ) driven by the Weather Research and Forecasting model (WRF). Three cases of
CMAQ simulations were conducted with emission data considering thermal power generation for the year
2010 and 2012, and without power plant emissions, using the meteorological field fixed to 2010.The
simulation for the year 2010 well agreed with observations. The emission increase caused higher air pollutant
concentrations around power plants, and the contribution of power plant emissions was up to 15 % of NO2
concentration in 2012.
Keywords: Thermal Power Plant, Air Quality Model, Meteorological Model, Nitrogen Dioxide, Sulfur
DioxideNational Center for Atmospheric Research. The
configuration of WRF and CMAQ used in thisstudy is shown in Table 1.

1. INTRODUCTION
The amount of thermal power generation has
significantly increased in Japan since the accidentof Fukushima Daiichi nuclear power station due tothe Great East Japan Earthquake in 2011. The ratioof nuclear power generation to the total power
generation in the major 10 companies in Japan was
29% in 2010. In 2012 after the Great East Japan
Earthquake, it decreased in 2%. The ratio ofthermal power generation rapidly increased from
62% to 89%. As the power generation in the Kinki
district of Japan depended heavily on nuclear
power generation, the ratio of thermal power
generation increased from 46% to 80%[1]. It
resulted in the increase in emissions of air
pollutants. It is important to assess air quality by
the increased air pollutants in the view of theenvironmental conservation.
In this study, the simulations of air quality by
the increased air pollutants from thermal power
generation after the Great East Japan Earthquake
were carried out in the Kinki district by using air
quality model.

Table 1 WRF/CMAQ configurations
Parameter
WRF
VersionInitial and boundaryLand use
Horizontal grid number
Vertical grid number
Explicit moisture
Cumulus

YSU scheme
Noah land-surface model
RRTM and Dudhia
Setting

Horizontal grid number

76×76 (D1), 92×104 (D2), 76×76(D3)

Initial and boundary

M ade from M OZART-4
INTEX-B，JATOP(vehiecle)，OPRF(ship)，
EAGrid2010-JAPANBaseline emission

2.1 Outline of Model

ARW 3.5.1
NCEP FNL, M SM -GPV, RTG-SST-HR
USGS 24-category data
98×88 (D1), 108×120 (D2), 92×92(D3)
30（surface-100 hPa layer）
WSM -6Kain-Fritsch

PBL
Surface layer
Radiation
Parameter
CM AQ
VersionHorizontal/verticaladvection
Horizontal/vertical
diffusion
Photolysis calculation
Gas phase chemistry
Aerosol

2. CALCULATION CONDITIONS

Setting

5.0.2

WRF-based scheme
M ultiscale/ACM 2
CCTM in-line calculation
CB05
AERO 6

Three simulations were carried out against
varying emissions from power generation; use of
the emissions in 2010 (2010case); use of theemissions in 2012 (2012case) and not considering
the emissions from power generation (base).
The boundary conditions of CMAQ were set tothe calculations by Model for Ozone and Related
Chemical Tracers version 4 (MOZART-4).

The air quality was simulated by Community
Multiscale Air Quality system (CMAQ5.02),
which was developed by United States
Environmental
Protection
Agency.
Themeteorological data was provided from thesimulation by Weather Research Forecastingmodel (WRF3.51), which was developed by578

GEOMATE- Bangkok, Nov. 14-16, 2016

each power stations in 2010 and 2011. In the
Kainan oil power station, the emissions extremely
increased in 2012, because the Kainan oil power
station was the peak load electricity source. On theother hand, the emissions in the Kobe coal power
station in 2012 were almost same as 2010, because
the Kobe coal power station was used as the baseload electricity source.

2.2 Calculation Period and domain

Maizuru
KainanKobe

Fig. 1 Location of thermal power plants in D3
Meteorology simulations were conducted byusing WRF from April 2010 to March 2011 withan initial spin-up period of 22-31 March 2010. The
calculation domains include domain 1 (D1)covering a wide area of Northeast Asia, domain 2(D2) covering almost the entire area of Japan anddomain 3(D3) covering the Kinki district. The
horizontal resolutions and the number of grid cells
are 64 km and 76 × 76 for D1, 16 km and 92 × 104
for D2, and 4 km and 76 × 76 for D3, respectively.
The vertical layers consist of 30 sigma-pressure
coordinated layers from the surface to 100 hPa
with the middle height of the first layer being
approximately 28 m. Figure 1 shows the domain 3
and the location of the thermal power plants.

Fig.2 NO X and SO 2 emissions from thermal plantin 2010 and 2012

3．RESULTS
3.1

Comparison between
observations

simulations

andThe simulations in WRF and CMAQ and theobservations were compared by several statisticalindexes, which were correlation confident (R),
Mean Absolute Error (MAE), Mean Bias Error(MBE) and Index of Agreement (IA). Table 2
shows the statistical indexes of mean daily
temperature, specific humidity and wind speed on
April, July, October and January in 2010 Japanfiscal year at several monitoring stations in Osaka
prefecture simulated by WRF. The criteria of thestatistical indexes, which were MBE ≦ ± 0.5oC,
MAE≦2oC, IA ≧0.8 for temperature, MBE≦±1
g/ kg, MAE≦2 g/kg, IA≧0.6 for mixing ratio andMBE≦±0.5 m/s, RMSE≦ 2 m/s, IA≧0.6 forwind speed, was proposed by Emery[5]. Except
for MBE of temperature on January, temperature,mixing ratio and wind speed satisfied the criteria.
Figure 3 shows the mean daily variations oftemperature, mixing ratio and wind speed in 2010
Japan fiscal year at several monitoring stations in
Osaka prefecture. The simulations well captured
the observations. Table 3 shows the statisticalindexes of mean daily NO 2 , SO 2 PM 2.5 and O 3 in2010 Japan fiscal year at several monitoring

2.3 Emissions from thermal power stations
The emissions of NOx, SO 2 , PM 10 , PM 2.5 ,NH 3 from thermal power stations shown in Figure
1 were estimated. The emissions of NOx, SO 2
from the thermal power stations in the Kansai
Electric Power Co. were given from the electric
power generation performance in 2014[2]. The
emissions of PM 10 , PM 2.5 , NH 3 were estimatedfrom the emission database of EAGrid2010JAPAN [3] in 2010 and fuel consumption in 2010
and 2012. The ratio of oil consumption, LNG
consumption and coal consumption in 2010 and2012 is 4.19, 1.56, and 1.15, respectively. The
emissions from the thermal power stations in othercompanies were predicted from the electric power
generation ratio in the Kansai Electric Power Co.
and other companies. The hourly variation of theemissions each month were considered from thehourly variation of the electric power generation in2012[4].
Figure 2 shows the emissions of NOx and SO 2
579

GEOMATE- Bangkok, Nov. 14-16, 2016

stations in Osaka prefecture simulated by CMAQ.
Figure 4 shows the mean daily variations of NO 2 ,SO 2 PM 2.5 and O 3 concentration in 2010 Japanfiscal year at several monitoring stations in Osaka
prefecture. The simulations well captured theobservations.

Fig.3 mean daily variations of temperature, mixing
ratio and wind speed in 2010 Japan fiscal year
Table 3 Statistical indexes of NO 2 , SO 2 PM 2.5 andO 3 concentration in Osaka prefecture
Statistic
NO 2
SO 2
PM 2.5O3
number
358
363
358
360
Obs.ave
22.03
4.77 19.21 49.82
Sim.ave21.80
3.05 14.27 49.17
R
0.85
0.65
0.88
0.84
MBE
-0.23 -1.72 -4.95 -0.66
MAE
4.04
2.00
5.70
7.79
IA
0.91
0.69
0.89
0.90

Table 2 Statistical indexes of temperature,mixing ratio and wind speed in Osaka prefecture
Temperature (2010JFY)
Statistic
Apr.
Jul.
Oct.
Jan.
number
720
744
744
744
Obs.ave
13.56 27.91 19.91
4.36
Sim.ave13.15 28.33 20.37
3.29
R
0.96
0.89
0.95
0.90
MBE
--0.420.42
0.47 -1.07
MAE
0.94
1.00
0.84
1.25
IA
0.97
0.93
0.97
0.90
Mixing ratio (2010JFY)
number
720
744
744
744
Obs.ave5.66 16.42
9.30
2.73
Sim.ave5.73 16.39
9.65
2.97
R
0.95
0.65
0.93
0.81
MBE
0.08 -0.03
0.36
0.24
MAE
0.53
0.86
0.71
0.35
IA
0.97
0.81
0.96
0.86
Wind Speed (2010JFY)
number
720
743
744
744
Obs.ave
2.63
2.54
2.11
3.08
Sim.ave
2.76
2.82
2.56
2.72
R
0.69
0.68
0.65
0.70
MBE
0.12
0.28
0.45 -0.35
MAE
1.28
1.26
1.09
1.48
IA
0.82
0.81
0.77
0.82

Fig.4 mean daily variations of NO 2 , SO 2 PM 2.5
and O 3 concentration in 2010 Japan fiscal year
3.2 Contribution to Air Quality by emissions
from thermal power stations
The detail analysis was performed at theneighborhood of Kainan thermal power station and
the Kobe thermal power station in which thepollutant emissions extremely increased in 2012
Figure 5 shows the NO 2 concentration in 2012, the580

GEOMATE- Bangkok, Nov. 14-16, 2016

contribution to NO 2 concentration by theemissions from thermal power station in 2010(2010-base) and the contribution to NO 2
concentration by the increment emissions fromthermal power station in 2012 (2012-2010) at the
Kainan and Kobe. The mean NO 2 concentration,
the mean contribution NO 2 concentration and themean contribution rate was 6.2ppb, 0.8ppb and13.4% at the Kainan and 11.6ppm, 0.2ppb and2.0% at the Kobe, respectively. NO 2 concentrationin the maximum contribution by the emission fromthe thermal power station was 24.1ppb at the
Kainan on 6 December and 20.5ppm at the Kobeon 13 March, respectively. The contribution NO 2
concentration and the contribution rate on theabove day was 3.7ppb and 15.3% at the Kainan
and 1.1ppb and 5.1% at the Kobe, respectively.

concentration by the increment emissions fromthermal power station in 2012 (2012-2010) at the
Kainan and Kobe. The mean SO 2 concentration,

the mean contribution SO 2 concentration and themean contribution rate was 3.3ppb, 0.9ppb and26.0% at

Fig.5 Time series of NO 2 concentration and thecontribution to NO 2 from thermal power station at
Kainan (top) and Kobe (bottom).
Figure 6 shows the distribution of NO 2
concentration in D3 and D1, and the distribution of
the incremental NO 2 concentration at 7 JST of 1
July, when the contribution to NO 2 concentrationby the increment emissions from thermal power
station at Kainan was maximum by the simulation.
The transboundary pollution of NO 2 didn’t
occurred in the simulation in D3. In the coast areas
and the industry areas, the high NO 2 concentration
occurred in the simulation in D1, because the main
emissions of NO 2 was from factory, vehicles andvessels. The remarkable increase of NO 2
concentration occurred at the Kainan and at the
Maizuru faced to Japan Sea. The increase of NO 2
concentration at the Kobe was small compared at
the Kainan. The spread of NO 2 concentration fromthe power stations was limited.
Figure 7 shows the SO 2 concentration in 2012,
the contribution to SO 2 concentration by theemissions from thermal power station in 2010(2010-base) and the contribution to SO 2

Maizuru

Kobe

KainanFig.6 Distribution of NO 2 concentration in D3(Top) and D1 (Middle), and the distribution of theincremental NO 2 concentration (Bottom) at 7 JSTof 1 July
the Kainan and 4.0ppm, 0.2ppb and 3.7% at theKobe, respectively. SO 2 concentration in the581

GEOMATE- Bangkok, Nov. 14-16, 2016

concentration by the increment emissions fromthermal power station in 2012 (2012-2010) at the
Kainan and Kobe. The maximum O 3 concentrationand the mean contribution O 3 concentration was
53ppb and -0.9ppb at the Kainan and 99ppm and 0.08ppb at the Kobe, respectively. The
contribution O 3 concentration by the emission
from the thermal

maximum contribution by the emission from thethermal power station was 13.3ppb at the Kainan
on 6 December and 12.0ppm at the Kobe on 5 May,
respectively. The contribution SO 2 concentrationand the contribution rate on the above day was
4.2ppb and 32.0% at the Kainan and 0.6ppb and5.4% at the Kobe, respectively.
Figure 8 shows the distribution of SO 2
concentration in D3 and D1, and the distribution of
the incremental SO 2 concentration at 7 JST of 5
May, when the contribution to SO 2 concentrationby the increment emissions from thermal power
station at Kainan was maximum by the simulation.
The transboundary pollution of SO 2 didn’t
occurred in the simulation in D3. In the coast areas
and the industrial areas, the high SO 2
concentration occurred in the simulation in D1,
because the main emissions of SO 2 was fromfactory and vessels. The remarkable increase of
SO 2 concentration occurred at the Kainan. The
increase of SO 2 concentration at the almost
thermalpower
stations
occurred.
SO 2
concentration diffused in the wide area.
The contribution rate of NO 2 and SO 2
concentration by the emission from the thermal
power station was almost same for the mean
concentration and for the maximum concentration.
These results suggested that the difference of thecontribution concentration occurred from thedifference of the meteorological conditions. SO 2
concentration relatively diffused in the wide area
compared with NO 2 concentration, because NO 2
was more reactive chemicals in the atmosphere.

Maizuru

Kobe

Kainannominal charges based on the agreement of theauthors.
Accepted papers which were notpresented
Fig.8 Distribution of SO 2 concentration in D3(Top) and D1 (Middle), and the distribution of theincremental SO 2 concentration (Bottom) at 7 JSTof 5 May
power station became negative because of NO
titration.
Figure 10 shows the distribution of O 3
concentration in D3 and D1, and the distribution of
the incremental O 3 concentration at 18 JST of 2
August, when O 3 concentration became maximumat Kainan by the simulation. O 3 concentration in
the simulation in D1 became high in inland thatFig.7 Time series of SO 2 concentration and thecontribution to SO 2 from thermal power station at
Kainan (top) and Kobe (bottom).
Figure 9 shows the O 3 concentration in 2012,
the contribution to O 3 concentration by theemissions from thermal power station in 2010(2010-base) and the contribution to O 3582

GEOMATE- Bangkok, Nov. 14-16, 2016

was apart from urban area with much air pollution
emissions. Due to the increment of emissions fromthermal power stations, O 3 concentration increasedin the wide area of D1 except for the neighborhoodof the thermal power stations because of NO
titration.

4. CONCLUSION
The simulations of air quality by the increased air
pollutants from thermal power generation afterFig.10 Distribution of O 3 concentration in D3(Top) and D1 (Middle), and the distribution of theincremental O 3 concentration (Bottom) at 18 JSTof 2 August
the Great East Japan Earthqu ake were carried out
in the Kinki district by using WRF/CMAQ. The
emissions of NOx and SO 2 from the thermal
power stations after the Great East Japan
Earthquake heavily increased. The simulations
showed that (1) the remarkable increment of air
pollution concentration of NO 2 and SO 2 waslimited in the neighborhood of the thermal power
stations, (2) the increment of O 3 concentration was
widely spread but O 3 concentration in theneighborhood of the thermal power stations
decreased because of NO titration.

Fig.9 Time series of O 3 concentration and thecontribution to O 3 from thermal power station at
Kainan (top) and Kobe (bottom).

4. REFERENCES[1] URL:http://www.fepc.or.jp/library/pamphlet/z
umenshu/[2] URL:http://www.kepco.co.jp/sustainability/kankyou/report/data/karyoku01.html[3] Fukui T., Kokuryo K., and et. al., Updating
EAGrid2000-Japan emissions inventory basedon the recent emission trends, Japan Society
for Atmospheric Environment, 49, 2014,
pp.117-125 (in Japanease).
[4] URL:http://www.kepco.co.jp/energy_supply/supply/
denkiyoho/download.html.

[5] Emery C., Tai E. and Yarwood G.: Enhanced
meteorological modeling and performance
evaluation for two Texas ozone episodes.
Prepared for The Texas Natural Resource
Conservation Commission 12118 Park 35
Circle Austin, Texas 78753, 2001.

583

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

OCEAN DECONTAMINATION: HIGH ABILITY REMOVAL METHODTO RADIOACTIVE CESIUM FROM OCEAN SLUDGE BY USINGMICRO BUBBLES AND ACTIVATING MICROORGANISMS
Kyoichi OKAMOTO1, Takeshi TOYAMA2 and Tomoe KOMORIYA3
1

Department of Oceanic Architecture and Engineering, CST, Nihon University, Japan;
2
Department of Materials and Applied Chemistry, CST, Nihon University, Japan2
Department of Sustainable Engineering, CIT, Nihon University, Japan

ABSTRACTThe Fukushima nuclear accident of March 11, 2011, soil and water had been contaminated by radioactive
cesium. Moreover, radioactive cesium was found in the ocean sludge in Tokyo Bay flowing from rivers. Cesium
which is adsorbed to the sludge cannot be easily removed. One of the authors developed decomposition andpurification system, a circulation-type system by micro bubbles, that is, by creating aerobic state, aerobic
bacteria are activated resulting to decomposition and purification of ocean sludge. Based on the hypothesis thatradioactive cesium is adsorbed on the surface of the sludge deposition. It is considered that cesium can be eluted
after decomposing the deposited sludge. Once the cesium is eluted in the water, it can fix to a mineral such aszeolite. Eluting and fixing cesium adsorbed on sludge takes so much time. In this case, the concept of removing
the left sludge by flocculation method and then followed by coagulating sedimentation method is studied. In thisstudy, our objects is to show the effectivity and efficiency of using flocculation and coagulating sedimentation inremoving radioactive cesium. As the results, we pointed out this method is very good.
Keywords: Ocean Decontamination, Radioactive Cesium, Micro Bubbles, Microorganism Activator, Coagulant

INTRODUCTIONDECOMPOSITIONCIRCULATION TYPE

The Fukushima nuclear accident of March 11,
2011, soil and water had been contaminated byradioactive cesium. Moreover, radioactive cesium
was found in the ocean sludge in Tokyo Bay flowing
from rivers. Cesium which is adsorbed to the sludge
cannot be easily removed.[1] One of the authors
developed decomposition and purification system, acirculation-type system by micro bubbles, that is, bycreating aerobic state, aerobic bacteria are activatedresulting to decomposition and purification of ocean
sludge. [2]Based on the hypothesis that radioactive cesium
is adsorbed on the surface of the sludge deposition,
it is considered that radioactive can be eluted after
decomposing the deposited sludge. Once the cesium
is eluted in the water, it can fix to a mineral such aszeolite. [3] But eluting and fixing cesium adsorbed
on sludge takes so much time. In this case, theconcept of removing the left sludge by flocculation
method and then followed by coagulating
sedimentation method is studied.
In this study, our object is to show the effectivity
and efficiency of using flocculation and coagulating
sedimentation in removing radioactive cesium.

SYSTEM

WITH

It is very important to reduce sedimentary sludge
in the ocean. Plans to reduce the sludge are usually
dreading or sand covering. Dredging is a simple way
and aims to cut off the sludge. But after cutting off,
treating the dredged sludge takes much more timeand, of course, cost. Sand covering, in general, gives
a big
wave

current

Sea Bottom

Micro-bubble Generating

Fig. 1
584

Pump

Experimental Tank

Purification System of Circulation Type.

GEOMATE- Bangkok, Nov. 14-16, 2016

which is to use flocculation method against to therest of decomposed sludge, and then make theprecipitation of cesium and sludge, shown in Fig. 4.

load to living organisms and the ecological system.
So that, a more efficient way is needed to reduce
the sludge while not imparting environmental load in
the local sea area. Here, attention was paid to microbubble technology for application to the purificationof the sludge. The important point in this technique
is to activate the bacteria existing in the area by
micro-bubbles. Micro-bubbles (that is MB) canchange conditions into an aerobic state. [4], [5] If the
bubbling stops, the situation changes into anaerobic
state, according to recent research. So, we selected amethod for decomposing the sludge by
microorganisms.
One of the authors had developed thedecomposition system for ocean sludge withcirculation type by micro-bubbles, shown in Fig.1,
which decompose and purification sludge byactivating the aerobic bacteria, after creating an
aerobic state by micro-bubbles. [3]Rest of
Decomposition

Cs+

-

Cs+

-

-

-

-

Cs+
Cs+

-

Cs+

- Cs+

Cs+

-

Cs+

Cs+

Cs+ - Cs+
PAC
- Cs+ PAC Cs＋- Cs＋ ＋＋-

Cs+

Coagulating
Sedimentation

＋

Cs＋

Cs＋

PAC＋

Experimental System
The experimental devices consist of two parts,
shown in Fig. 5. The water circulates through twotanks. In one tank (Width40xLength28x Hight28cm),
micro-bubbles are generated. The micro-bubbles
have micro-size diameter and high solubility. Thismeans the water with high concentration ofdissolved oxygen circulates through these tanks. The
other
part
is
the
experimentaltank(W60xL29xH35cm). We used sea-water 30(litter)and sludge 1(kg). Here, a micro-bubble generator isbased on [4], [5] and the flow rate is 900 (litter/hour).
The flow rate of water pumps connected each tanks
are 300 (litter/hour). A cooler for water tank was set
at side of the tank for generating microbubbles, forthe purpose of setting water temperature 30 degree
centigrade.

- Cs+ -

-

Cs+

PAC＋

REMOVAL EXPERIMENTS FOR CESIUM

Cs+

Cs+

-

Cs＋

- Cs＋PAC-

Fig. 4 Mechanism on Coagulating Sedimentation
by PAC, to the left of Decomposition Sludge with
Cesium．

- Cs+

Cs+

-

-

Fig. 2

Cs+

-

Cs+

Cs+

- Cs+

- Cs＋ -

Adsorption

Cesium- Cs+ -

Cs+

Cs＋

Cs＋

In general, ocean sludge has a negative charge.
When cesium with a positive charge flows fromriver, sludge was adsorbed cesium, shown in Fig. 2.
So that, sludge adsorbed cesium cannot eliminate byusual way.

-

- Cs+

Cs+

MECHANISM ON REMOVING CESIUM

SludgeFlocculation-

Mechanism on Adsorption of Cesium.

Cooler

ZeoliteHere, we have a way by using of thedecomposition system for ocean sludge withcirculation type. After decomposition of the sludge
adsorbed cesium by our system, cesium is eluted
into water, shown in Fig. 3. That is our hypothesis.
Adsorption
Cs+
-

Cs+

-

Cs+

- Cs+

Cs+

- Cs+ -

Cs+

Cs+

-

Cs+
Cs+

Pump for Circulation

Cs+

Cs+

Cs+

MB Generator

Cs+

Cs+

-

Cs+

-

ElutionDecomposition

Cs+

Cs+

FixingFix

Fig. 5 Experimental System for Removal of
Cesium.

bｙ UsualTechnology
ex. Zeolite

Not elute
In the adsorbed
state

Experimental Procedure
We had caught the sludge and the sea water at
Funabashi Port in Chiba Prefecture in JAPAN, as
shown in Fig.4 and 5. Here, we had picked up thesludge under 10cm from seabed before sampling as
experimental procedure, because we have to cut theFig. 3 Mechanism on Fixing of Cesium after
Elution.
But, it takes much time to elute and fix all of
cesium adsorbed on sludge. Now, we have new idea
585

GEOMATE- Bangkok, Nov. 14-16, 2016

Procedure for Solid Measurements

initial value of cesium in the sludge, from [3].
We used the cesium chloride before 24 hours ofstarting time and the concentration of cesium ion is
100 (ppm).
After setting the decomposition system withcirculation type by micro-bubbles, experiment starts
at the same time of generating micro-bubble device
and also the zeolites were set in the tank. The zeolite
is the composed type and the pore size is 4A typethat is 0.4[nm]. Diameter of cesium is generally
0.338[nm], so that this diameter is much closer.

As procedure of solid phase, experimental tanks areprepared for each measurement time; 0, 24, 48, 72,
96, 120. When the objective tank after worked
system is stopped the system, coagulant was put andmixing. After this, water quality and cesium aremeasured. After filtration and dry, cesium in solid
was analyzed by the energy dispersion type X-ray
analysis device (EDX). Here, our used coagulant is
PAC (Poly Aluminum Chloride) and the solution
concentration is 10%. [6]
Start

Tokyo

Putting
PAC

Generate MB

Yokohama-24h

Set Sea Water, Sludge
and Cesium

0h

6h

24h

48h

72h

Filtration
and Dry

Set Activator of
Microorganisms

Fig. 8 Experimental Procedure of for Solid
Measurements (For Example: 72hours Case).

Experimental Conditions
Experimental conditions are the concentration of
PAC according to the putting time, shown in Table 1.
The another cases are PAC concentration is 500 ppm
at 0, 24 hours, 25 ppm at 96 hours and 10 ppm at
120 hours.

Fig. 6 Catching Point of Sludge and Sea Water at
Funabashi Port in Tokyo Bay.
After 6 hours later, the microorganism activator
was put in the experimental tank. Main staff of theactivator is Kelp and including nutrients and someenzyme. Our used activator is reported to show
effective results in purification for grease trap.
Dissolved oxygen (DO), water temperature andpH are measured by using of multi-parameter water
quality meter. Ammonium nitrogen (NH4-N), total
nitrogen (T-N) are measured by using of digitalwater-analyzer by digital “Packtest”, by water
filtered after sampling in experimental tank.

Table 1 Experimental Conditions
at 48hPAC (ppm)
600
400
200

at 60hPAC (ppm)
200
100
75

at 72hPAC (ppm)
100
75
50

Procedure for Liquid Measurements

RESULTS AND CONSIDERATION

Measurements for liquid phase did each 6 hours
until 12 hours, and then every 12 hours to120 hours,
shown in Fig.7.

Results on Water Quality
Water Temperature, pH, DO (Dissolved Oxygen)and H2S (Hydrogen Sulfide)

Start
Generate MB-24h
Set Sea Water, Sludge
and Cesium

0h

6h

24h

48h

72h

96h

As the results of environmental conditions; watertemperature, pH and DO, water temperature wasalmost 30 centigrade degrees because of using watercooler. pH showed the values of 7.5 to 8.7. DO issaturation state by the values of 7.5 to 8.0. These
results are almost same as ordinary results. [3]H2S decreased rapidly by 24 hours by the supplyof O2 of microbubble device and then became zero

120h

Set Activator of
Microorganisms

Fig. 7 Experimental
Measurements.

Procedurefor

Liquid586

GEOMATE- Bangkok, Nov. 14-16, 2016at 48 hours. This is also the same as ordinary results.
[3]

Concentration of PAC (ppm)

600

DIN (Total Inorganic Nitrogen), T-N (Total
Nitrogen) and Cesium in Water
DIN and T-N as purification items decreased
well by the denitrification. T-N is reduced to 82%.
This is also the same as ordinary results. [7], [8]Cesium in liquid was analyzed by the ironchromatography, and the initial value for cesium is50.06 (ppm) but the value by 12 hours is notdetective, so that reduction rate of cesium was 100%.

DIN, TN (mg/L)
Cs in Water (10*mg/L)

6500ｐｐｍ500

400ｐｐｍ

400
300
200

75ｐｐｍ
100

10ｐｐｍ
25ｐｐｍ

100ｐｐｍ
0
24

0

48

72

96

120

Time (hour)Fig. 10
Suitable
Concentrationof
PAC
According to Working Time in Experimental
System.

DIN (mg/L)
TN (mg/L)

5

ADDITIONAL EXPERIMENT

Cs in Water (10* mg/L)

4

We get the relationship of most suitableconcentration of PAC according to working time ofthe system. But we need the changes of cesium
before and after working system.
Now we executed the additional experiments forthe purpose of getting of the amount of cesium
before and after working system. The experiments
are made in cases of 48 hours and 72 hours, and ascomparative target is the case of 0 hour.

3
2
1
0
0

24

48

72

96

120

Time (hour)Fig. 9Changes on DIN, TN and Cesium in Water.

Water Quality

Most Suitable Concentration of PAC

The results for water quality are shown Fig. 11.
DIN and T-N are very good decrease by the
denitrification. This is also the same as our ordinary
results.

We are going to find the most suitableconcentration of coagulant (PAC) depended on theworking time, by putting the coagulant withconcentration according to experimental condition.
By focusing on the T-N as the index of Table 2,
the relationship of minimum concentration of PAC
and working time of the system was shown in
Fig.10. Here, the results of T-N were N.D. in caseof 0, 24, 96 and 120 hours.
When it’s the concentration of PAC of the upper
part of a solid line in Fig.10, cesium of solid phaseindicates the maximum purification performance andcoagulating sedimentation is executed. Therefore,
most suitable working time of the system is 48 to 60
hours, so that the reducing working time can besupposed and expected.

3.5
DIN (mg/L)
TN (mg/L)

DIN, TN (mg/L)

3
2.5
2

1.5
1

0.5
0
0

12

24

36

48

60

72

Time (hour)Fig. 11
Changes on DIN, TN in Additional
Experiment.

Table 2 Experimental Results.
Cesium in Water
at 48hPAC (ppm) TN/TN0
600
ND
400
ND
200
0.24

at 60hPAC (ppm) TN/TN0
200
ND
100
ND
75
0.22

at 72hPAC (ppm) TN/TN0
100
ND
75
ND
50
0.20

Cesium in liquid was analyzed by atomic
absorption spectrometry, and the value for cesium is
not zero at 48 hours and 72 hours, shown in Fig.12.
This is caused by weak circulation of experiment,
and this analytical way is very good precision. But
this is not big problem. We can understand the587

GEOMATE- Bangkok, Nov. 14-16, 2016

eluting cesium from sludge and have another way;adsorption cloth and other for cesium adsorption andfixing.
The difference is small in the cesium amount
before and after investment of PAC, in case of at 48
and 72 hours.

assumed this concentration of PAC shows the
maximum purification efficiency in water and isvery good.
(3) It seemed the most suitable time for working theexperimental system is 48 to 60 hours, so thatthe working time of system can be reduced.
We also executed the additional experiments
for the purpose of getting of the amount of cesium
before and after working system, in case of 48 hours
and 72 hours, and 0 hour as comparative target. As
the results are

Cs in Water (10*mg/L)

6.00

72 hour

5.00

48 hour

4.00
3.00

22.4

2.00

22.4（200ppm）
21.1（400ppm）
20.4（600ppm）

1.00
0.00

(4) Cesium in water after stopping of the system
and investment of PAC in case of 48 and 72
hours, is almost the same as the quality before
investment of PAC.

17.0（ 50ppm）
14.0（ 75ppm）
14.2（100ppm）

17.5

0

12

24

36

48

60

72

Time(hour)

Changes on Cesium in Water before andFig. 12after investment of PAC.

ACKNOWLEDGEMENTS
The authors would like to express sincere thanks
for executing the experiments to Takahiro
ARAKAWA and Taiyo KOKATSU, Students of
Nihon University, in Japan.
This work was supported by KAKENHI (Grantin-Aid for Scientific Research (C-26420831)).

About Safety to all Ecosystem by this System
Coagulant used in this experimental system is PAC(Poly Aluminum Chloride) and the solution
concentration is 10%. [6] This is basically safe
according to appropriate usage. It’s supposed PAC
is certainly the substance which isn't diffused aroundthe environment. It seems it is no problem as far asclear layer at the top of liquid is discharged, because
coagulant power is strong. Toxicity to short-neckedclam and laver (Medial Tolerance Limit) is
10000(ppm)/48(hour) from safety data sheets (SDS).
[9] Even if PAC we added doesn't have an influenceon the environment. PAC hydrolyzes with passage
in time and changes into a stable aluminumhydroxide.

REFERENCES[1] Tokyo News Paper: Radioactivity pollution in
the Tokyo Bay, ”A nuclear power plant, farconvergence”, March 10, 2015.
http://www.tokyo- np.co.jp/feature/ tohokujisin/archive/fouryears/150310_1.html[2] Okamoto K, Hotta K, Toyama T and Kohno H,“Purification System of Ocean Sludge by
Activating Microorganisms”, International
Journal of GEOMATE (Geotec, Const. Mat. &Env.), Vol.6, No.1, pp.791-795, March, 2014.
[3] Okamoto K and Toyama T: Ocean
Decontamination: Removal of Radioactive
Cesium from Ocean Sludge by Using Microbubbles and Activating Microorganisms, The
Japan Society of Naval Architects and Ocean
Engineers, May, 2014.
[4] Imai T, Shioshige K, Ukita M, Sekine M,
Higuchi T, Fukagawa K and Fujisato T,“Development of Device for Dissolving High
Concentration Gas for Purification at bottom
layer around enclosed water area like Dam
Lake”, Proc. 39th Forum on Environmental
Engineering of Civil Engineering Society,pp.10-12, 2002.

CONCLUSION
We carried out the removing cesium in solid
phase for experimental system which is to use
flocculation method against to the rest ofdecomposition sludge, after the elution and fixing
cesium by setting zeolite in the decomposition
system with circulation type by micro-bubble andactivating microorganisms.
(1) From the results by measurements for water
qualification, T-N decreases maximum 82.0%.
Cesium in water was fixed 100% and becamezero by 12 hours of the experiment.
(2) We pointed out the relationship of mostsuitable concentration of PAC according toworking time in experimental system. It
588

GEOMATE- Bangkok, Nov. 14-16, 2016[5] Matsuo K, Maeda K, Ohnari H, Tsunami Y and
Ohnari H, “Water Purification of a Dam Lake
Using Micro Bubble Technology”, Progress inMultiphase Flow Research I, pp.279-286, 2006.
[6] Okamoto K and Hotta K, “Purification System
of Ocean Sludge by Using Coagulants andActivating Microorganisms”, International
Journal of GEOMATE (Geotec, Const. Mat. &Env.), Vol.4, No.2, pp.574-579, June, 2013.
[7] Sone T, Yamashita K and Okamoto K,“Identification of Microorganism in Purification
Process for Ocean Sludge by using Purification

System with Circulation Type”, Japan
Association for Coastal Zone Studies, July 2014.
[8] Okamoto K and Toyama T, ”Ocean
Decontamination: High Efficiency Removal
Method of Radioactive Cesium from Ocean
Sludge by Using Microbubbles and Activating
Microorganisms”, The Japan Society of Naval
Architects and Ocean Engineers, May, 2016.
[9] Poly Aluminum Chloride (PAC), UPC-0100,
Safety Data Sheets (SDS), January, 2014 (inJapanese).

589

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EFFECT OF ADDITION OF BACTERIA ON THE REMOVAL OF
RADIOACTIVE CESIUM FROM OCEAN SLUDGE IN A
CIRCULATION TYPE PURIFICATION SYSTEM
Tomoe KOMORIYA1, Kyoichi OKAMOTO2 and Takashi TOYAMA2
1
CIT, Nihon University, JAPAN; 2CST, Nihon University, JAPANABSTRACT
Following the Fukushima nuclear accident of March 11, 2011, soil and water were contaminated byradioactive cesium. Moreover, radioactive cesium was found in the ocean sludge in Tokyo Bay, carried by rivers
flowing into the bay. The cesium adsorbed in the sludge cannot easily be removed. The objective of this study
was to investigate the effect of the addition of bacteria to the micro-bubble circulation system on the efficient
removal of radioactive cesium from ocean sludge. One of the authors has developed an ocean sludge
decomposition system employing circulation of micro-bubbles. Model sludge was prepared using seawater, sea
sludge, and cesium chloride. Bacteria were added to the system after 24 h. Dried tangle extract was added as anutrient at 24 h and 36 h. The decomposition experiment was carried out for 120 h. The circulation of microbubbles created an aerobic state that activated aerobic bacteria, facilitating decomposition and purification of thesludge. Thus, decomposition of the deposited sludge using our system renders the elution of the radioactive
cesium possible. If the cesium is eluted in the water, we can fix it using existing technology such as zeolites. We
identified and isolated the most useful bacteria for sludge decomposition. Effects on purification seem to begreatest when additional bacteria are added directly to the process. The methodology proposed is expected tofacilitate decomposition of sludge and removal of radioactive cesium from the environment.
Keywords: Decontamination, Radioactive Cesium, Ocean Sludge, Micro-bubble, Bacteriasuch as zeolites [4]. We would thus be able todecontaminate the sediment.
To remove radioactive cesium effectively, we
carried out isolation and identification of bacteria to
break down sludge in the micro-bubble circulation
system, and investigated the effect of the addition of
microorganisms to the decomposition system on theefficient removal of radioactive cesium from ocean
sludge.

INTRODUCTION
The 2011 accident at the Fukushima Daiichi
nuclear power station has resulted in radioactive
cesium contamination of soil and water. Radioactive
cesium has been detected in the sludge in Tokyo Bay[1], carried by rivers flowing into the bay.
It has been reported that cesium easily adsorbs tothe microscopic particles that constitute soil [1], [2].
Most cesium adsorbed by soil is difficult to removeby external factors, and remains present over longtimescales. Moreover, closed water areas such as
river and bay systems make it difficult to decompose
piling organic sludge. Therefore, radioactive cesium
is predicted to be deposited in the sediment of theseabed over time, expanding contamination into theocean. It is important to decontaminate this sediment.
One of the authors developed a decomposition
system for ocean sludge that employs the circulationof micro-bubbles. These create aerobic conditions
that activate
aerobic bacteria, facilitating
decomposition and purification of the sludge [3].
Accordingly, it is considered that radioactive cesium
can be eluted, after decomposition of the deposited
sludge by our system. If the cesium can be eluted inwater, it can then be fixed using existing technology

MATERIALS AND METHODS
Isolation and Identification of Bacteria to Break
Down Sludge in the Micro-Bubble Circulation
System
The micro-bubble circulation system consists of
two parts, shown in Fig. 1 (without zeolite). The
water circulates through two tanks. In one tank(length 40 × width 28 × height 28 cm), microbubbles are generated. The micro-bubbles havemicro-size diameter and high solubility. This meansthat water with a high concentration of dissolved
oxygen (DO) circulates through these tanks. The
experimental tank is 60 × 29 × 35 cm. We used 30 Lof seawater and 1 kg of sludge. The micro-bubble
generator was based on [6] and the flow rate was590

GEOMATE- Bangkok, Nov. 14-16, 2016

900 L/h. The flow rate of the water pumps connectedto each tank was 300 L/h. A cooler for the tank thatgenerates the micro-bubbles was set at 30 °C.

Fig. 1After 24 h, isolated bacteria (15 × 108 cell/mL)
and an activator [7] were added. An extract of kelp
was added to the experimental tank after 24 and 60 h.
After 0, 12, 48, 60, 72, 96, and 120 h, DO, watertemperature, and pH were measured by a multiparameter water quality meter. A digital pack test
(Kyoritsu Chemical-Check Lab. Corp., Japan) wasused to measure ammonium ions (as ammonium
nitrogen, NH 4 -N), nitrite ions (as nitrite nitrogen,NO 2 -N), nitrate ions (as nitrate nitrogen, NO 3 -N),
total nitrogen (T-N), and hydrogen sulfide (H 2 S).
The seawater was then filtered. We checked theshape of the sea sludge and the effect of cesium
removal with SEM-EDX, after 0 and 120 h.
Experimental conditions, including the amount ofkelp used, are shown in Table 1.

Circulation purification system

We took samples of sludge and seawater at
Funabashi Port in Chiba Prefecture in Japan. We
removed the first 10 cm of sludge from the seabed
before samples were taken.
After setting up the micro-bubble circulation
decomposition system, the experiment started as thegeneration of micro-bubbles began. After 0, 36, 72,
108, and 120 h, we sampled the sludge in theexperimental device.
Samples were diluted in distilled water and 10
μL of each sample was plated on standard agar
plates (Nissui Pharmaceutical Co., Ltd., Japan) andcultured at 30 °C for 3–5 days. Thereafter, abacterial colony was picked for isolation. The
isolated bacteria were identified by 16S rRNA genesequence analysis.

Table 1 Experimental conditions

Case 1Case 2
Case 3

Activator

Extract of
kelp

−
−
100 ppm

500 ppm−−

RESULTS AND DISCUSSION
Isolation and Identification of Bacteria
Water was sampled from the experimental device
after 0, 24, 72, and 120 h. Fifty-three strains ofbacteria were isolated from these samples andidentified by 16S rRNA gene sequence analysis. Six
species of bacteria were identified from samples at 0,36 and 72 h: Bacillus cereus, B. clausii, B.
licheniformis, B. thuringiensis, B. subtilis, andAlcaligenes fecalis.
B. clausii is alkaliphilic and produces highalkaline proteases. B. licheniformis is commonly
known to cause food poisoning and food spoilage. B.
subtilis is non-pathogenic. It can contaminate food,however seldom results in food poisoning. B. cereus
produces two types of toxins and is known to cause
food poisoning. B. thuringiensis is a close relative of
B. cereus. It produces insecticidal protein. Thus,
Bacillus species are almost ubiquitous in nature,
however, epidemiologically they are unsuitable touse for sludge decomposition.
Alcaligenes fecalis is commonly found in theenvironment in soil and in wastewater disposal
apparatus. A. fecalis has heterotrophic nitrification
and aerobic denitrification functions [8]. Thus, we
assumed that A. fecalis would be useful fordecomposition of deposited sludge. We carried out
experiments of decomposition of the deposited
sludge using our system (Fig. 1) with micro-bubbles
and A. fecalis.

Experiments on Decomposition of Deposited
Sludge for Cesium Removal
The system shown in Fig. 1 was used toexperimentally remove cesium from ocean sludge.
Model sludge containing cesium chloride waspoured into the system. One kg of sea-sludge and 30
L of seawater and cesium chloride was mixed andstirred for 24 h. The final concentration of cesium
was approximately 100 ppm.
Zeolites were placed in the tank (Fig. 2), microbubbles were generated, and the experiment began.
The micro-bubble generator again used a flow rateof 900 L/h. The flow rate of the water pumps
connected to each tank was 2400 L/h. The cooler
was set at 30 °C.

Fig. 2

Additiveamountof bacteria
1.5 × 108 cell
1.5 × 108 cell−

Zeolites set in experiment tank591

GEOMATE- Bangkok, Nov. 14-16, 2016

Experiments of Decomposition of Deposited
Sludge for Cesium Removal
Results of water temperature, pH, DO and H 2 S invarying environmental conditions
Figures 3–5 show the water temperature, pH, andDO respectively, resulting from the environmental
conditions of cases 1–3 (Table 1). Water temperatureis almost constant at about 30 °C after 6 h, as a resultof using a cooler for the experimental water tank.
The pH of case 1 is constant at about 9.5, and that ofcases 2 and 3 is constant at about 7.5 to 8.0: the pHof case 1 was higher than cases 2 and 3. The initialconcentrations of DO are 1.9 mg/L, 5.3 mg/L and5.6 mg/L in cases 1, 2, and 3 respectively. The DOof all cases is saturated at about 7.5 to 8.3 mg/L after
6 h, because the concentration of oxygen saturationis 8.11 mg/L-pure water.

Fig. 3Fig. 5

Change in DO concentration over time incase 1 (●), case 2 (∆), and case 3 (■).

H 2 S concentration is shown in Fig. 6. In all cases,
H 2 S decreases up to 24 h, before the addition of A.
fecalis to the experimental tank. A. fecalis is notinhibited by H 2 S.

Fig. 6

Change in water temperature over time incase 1 (●), case 2 (∆), and case 3 (■).

Change in H 2 S concentration over time incase 1 (●), case 2 (∆), and case 3 (■).

Results of NH 4 -N, NO 2 -N, NO 3 -N and T-N

Fig. 4

Figure 7 shows the change in NH 4 -N
(ammonium nitrogen) concentration in theexperimental tank over time. In all three cases (Table1), initial concentrations of NH 4 -N are
approximately 2.0 mg/L. In case 1, NH 4 -N
decreases; at 24 h it is no longer detected. The effectof A. fecalis on NH 4 -N concentration could not be
ascertained, because A. fecalis was added at 24 h. Incase 2, NH 4 -N decreases up to 48 h and thenremains constant at 0.2 mg/L. In case 3, NH 4 -N
decreases up to 24 h and then remains constant at 0.2
mg/L.

Change in water pH over time in case 1 (●),
case 2 (∆), and case 3 (■).

592

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 7

Change in NH 4 -N concentration over timein case 1 (●), case 2 (∆), and case 3 (■).

Fig. 9Figure 8 shows the change in NO 2 -N (nitrite
nitrogen) concentration in the experimental tankover time. In case 1, initial concentrations of NO 2 -N
are approximately 0.30 mg/L. Subsequently NO 2 -N
decreases up to 72 h, after which it is constant at
0.80 mg/L. In case 2, initial concentrations of NO 2 N are approximately 0.20 mg/L. NO 2 -N increases to
0.4 mg/L at 24 h before decreasing up to 96 h, andthen remains constant at 0.08 mg/L. In case 3, initialconcentrations of NO 2 -N are approximately 0.15
mg/L. Following this, NO 2 -N increases to 0.25 mg/L
at 24 h, then decreases up to 72 h after which it
remains constant at 0.01 mg/L.

Fig. 8Change in NO 3 -N concentration over timein case 1 (●), case 2 (∆), and case 3 (■).

In the initial 24 h, NH 4 -N concentration
decreased, and NO 2 -N and NO 3 -N concentrations
increased. It is assumed that the metabolism of thebacteria switched from denitrification to nitrification
because the experimental tank water changed fromanaerobic to aerobic conditions (Fig. 5). Therefore
we suppose that the nitrogen source shifts with NH 4 +→ NO 2 − → NO 3 − and that this trend of NO 3 -N
concentration was similar to NO 2 -N.
Dissolved inorganic nitrogen (DIN; NH 4 -N +NO 2 -N + NO 3 -N) shows 85 % and 90 % decreases
in cases 1 and 3 respectively (Fig. 10).

Fig. 10 Change in DIN concentration over time incase 1 (●), case 2 (∆), and case 3 (■).

Change in NO 2 -N concentration over timein case 1 (●), case 2 (∆), and case 3 (■).

The effect of the addition of A. fecalis on T-N(total nitrogen, and inorganic and organic nitrogen)concentration was investigated. After adding A.
fecalis to the water, T-N decreased and was
eventually no longer detected (Fig. 11).

Figure 9 shows the change in NO 3 -N (nitrate
nitrogen) concentration in the experimental tankover time. In case 1, initial concentrations of NO 3 -N
are approximately 4.3 mg/L. Subsequently NO 3 -N
decreases up to 96 h, after which it remains constant
at 0.9 mg/L. In case 2, initial concentrations areapproximately 2.0 mg/L. Following this, NO 3 -N
increases to 6.4 mg/L at 24 h, then decreases to 3.0
mg/L at 60 h before continuing to gradually decrease
to 1.5 mg/L at 120 h. In case 3, initial concentrations
are approximately 2.0 mg/L. NO 3 -N increases to 3.9
mg/L at 24 h, then decreases to 0.2 mg/L at 96 h.
NO 2 -N and NO 3 -N showed similar trends.

Observation of sludge shape
The shape of the sea sludge was observed using
SEM at 0 and 120 h in case 1. Fig. 12 shows theSEM photomicrographs at 0 (a), and 120 (b) h. At
120 h after exposure to micro-bubbles and bacteria,
the sludge particles had become smaller.

593

GEOMATE- Bangkok, Nov. 14-16, 2016

sludge was decomposed by micro-bubbles and A.
fecalis.

Fig. 11 Change in T-N concentration over time incase 1 (●), case 2 (∆), and case 3 (■).
Fig. 13 Change in Cs/Si over time in case 1.
(a)
CONCLUSION
We carried out elution and fixing of cesium by
micro-bubble and bacteria circulation decomposition
system for ocean sludge. From water quality
measurements, it can be concluded that after adding
A. fecalis, T-N decreases and then is no longer
detected. A. fecalis has a remarkable ability to treat
samples through denitrification.
The sea-sludge particles decrease in size throughtreatment with micro-bubbles and bacteria for 120 h.
We predict from the results shown here that thesystem proposed would greatly facilitate cesium
removal from ocean sludge.

(b)

ACKNOWLEDGEMENTS
The authors would like to express sincere thanks
to Takaaki Sone, Hiroki Kazama and So Miyamoto,students of Nihon University, in Japan, for executing
the experiments.
This work was supported by KAKENHI C26420831 (Grant-in-Aid for Scientific Research).
Fig. 12 SEM photomicrographs of sludge (a) before
experiment and (b) after 120 h in case 1.

REFERENCES[1] Soemori A, Shozugawa K, Nogawa N, Higaki S,
Matsuo M, “A change in the concentrations ofradioactive cesium in Tokyo-bay’s sediments
released by the Fukushima Dai-ichi Nuclear
Power Station accident”, Bunseki Kagaku, Vol.
62(12), 2013, pp. 1079–1086.

Results of cesium concentration
We measured cesium and silica in the sludge
using energy dispersive X-ray analysis (EDX). The
weight ratios of cesium to silica (Cs/Si) in dry
sludge are shown in Fig. 13. The cesium
decontamination ratio was calculated from the ratiosof cesium content at 0, 24, 72, and 120 h as
measured by EDX, using standard values for silica
after measuring the weight of the dried sludge. The
decontamination ratio obtained in case 1 was about
50 %.
It has been reported that cesium concentration inwater decreased abruptly after the beginning ofsimilar experiments [9]. Therefore, eluted cesium
from the sludge was adsorbed to zeolite, after the[2] He Q, Walling D E, “Interpreting particle size
effects in the adsorption of 137Cs andunsupported 210Pb by mineral soils andsediments”, J. Environ. Radioactivity, Vol. 30(2),1996, pp. 117–137.
[3] Okamoto K, Hotta K, Toyama T, Kohno H,“Purification system of ocean sludge byactivating microorganisms”, Int. J. of
GEOMATE, Vol. 6(1), 2014, pp. 791–795.
594

GEOMATE- Bangkok, Nov. 14-16, 2016[4] Okamoto K, Toyama T, “Decontamination ofradioactive cesium from ocean sludge by microbubble and microorganisms”, Int. J. of
GEOMATE, Vol. 9(1), 2015, pp. 1390–1394.
[5] Tsukada H, Takeda A, Hisamatsu S, Inaba J,“Concentration and specific activity offallout 137Cs in extracted and particle-size
fractions of cultivated soils”, J. Environ.
Radioactivity, Vol. 99(6), 2008, pp. 875–881.
[6] Matsuo K, Maeda K, Ohnari H, Tsunami Y,
Ohnari H, “Water purification of a dam lakeusing micro bubble technology”, Progress inMultiphase Flow Research, Vol. 1, 2006, pp.
279–286.

[7] Okamoto K, Hotta K, “Purification experimentson sedimentary sludge by microorganism
activator”, In Proc. of Pacific Congress on
Marine Science and Technology (PACON), 2008.
[8] Inamori Y, Wu X-L, Kimochi Y, “Characteristics
of N 2 O production by Alcaligenes faecalis underaerobic conditions”, Journal of Japan Society on
Water Environment, Vol. 22, 1999, pp. 904–909.
[9] Okamoto
K,ToyamaT,“Oceandecontamination: Removal
efficiency ofradioactive cesium from ocean sludge by using
micro bubbles and activating microorganisms”,
Int. J. of GEOMATE, Vol. 10(21), 2016, pp.
1924–1928.

595

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

A STUDY ON GROUND IMPROVEMENT TECNIQUE WITH IN-SITU
MICROOGANISMS ISOLATED FROM JAPAN
Rusutsu Ito1, Toshiro Hata1
Graduate School of Engineering, Toyama Prefecture University, Japan

1

ABSTRACT
Damage by a liquefaction phenomenon was a problem in recent years, and the Great East Japan Earthquake
occurred on March 11, 2011 in Japan, and liquefaction damage occurred frequently. There is also more adoptionof Paris agreement in COP21 of the end of last year, and correspondence to a global warming problem is also
desired reduction in greenhouse effect gas amount of emission in the construction field. Therefore consideredground improvement technology is necessary for the environment in Japan an earthquake-ridden country. So we
considered for practical use ground improvement techniques based microorganism. That is watched as newliquefaction countermeasure technology for reduction in cost and the point of view by which material andconstruction waste are reduction. In that ground improvement techniques, it is difficult to using specificmicrobes. So we aimed at Microbial carbonate precipitation using in-situ microorganism as the method to solve
this problem. We made solidify sand using isolated microorganisms in japan and Bacillus pasteurii thesolidification ability becomes clear. We measured to urease activity values of each microorganism. And wemaking of the test pieces, undrained cyclic triaxial test and acid decomposition for using CaCO3(0.5mol/L). And
we compared results. We understood two things from examination results. 1) The difference occurs toliquefaction strength by urease activity value. 2) The improvement effect of the liquefaction strength wasadmitted in Microbial carbonate precipitation using in-situ microorganisms in the spots selected by this research.

Keywords: Microbial carbonate precipitation, Urease activity, Undrained cyclic triaxial test, Liquefaction

INTRODUCTION

solidification by a calcium carbonate way in the past,
the case for which the microorganism from which
unlined clothing was separated already (alien
species) is used is seen much. But we can think theuse of a foreign microorganism is difficult frominfluence to a local ecosystem when making apractical use of. We aimed at utilization of a in situ
microorganism as the method to settle this problem
by this research.
The soil stabilization technology for whichmicroorganism solidification was used was made apractical use of with a final goal in Japanese wholeand Abashiri city, Hokkaido was selected as aJapanese northern end by this research. Imizu city,
Toyama was selected as the spot where thedangerous degree of the liquefaction is higher thanToyama-ken liquefaction map and isolation of aurase activated positive microorganism where eachspot lives in the ground and the liquefaction strength
improvement effect were inspected. We write down
the point that isolated a microbe in figure -1. We
also inspected about Bacillus pasteurii it's said thatthe solidified effect is clear where at the same timeand considered the validity of the suggestion
technology for which a in situ microorganism isused through comparison of a result.

The Great East Japan Earthquake occurred on
March 11, 2011. Many earthquakes also occur toothers in Japan which is an earthquake-ridden
country. The liquefaction phenomenon which occurs
with those earthquakes in recent years is a problem,
and it's said that they need liquefaction
countermeasure technology which corresponds tovarious cases. Correspondence to a global warming
problem is also desired reduction in greenhouseeffect gas amount of emission in the construction
field. So reduction in greenhouse effect gas amountof emission was expected more than conventional
technology by this research. And the calcium
carbonate way which is the soil stabilization
technology forwhich themicroorganism
solidification watched as new liquefaction
countermeasure technology was used wasconsidered for practical use.
BACKGROUND
To use the microorganism with the higher urase
activated value (U/L), when doing microorganism
596

GEOMATE- Bangkok, Nov. 14-16, 2016

MATERIALS AND METHODS

an isolated microbe from all over the soil in Imizu
city, Toyama, solidification ability solidified using
Bacillus pasteurii which became clear in this study.
We write down in list of examination cases table -1,
composition of the solidified liquid table -2, an
enzyme activity level of the microorganisms whichmeasured it using LCR meter in composition of thesolidification solution, table -3. We performed
repetition Undrained cyclic triaxial test (DA=5%) of
the soil with the test specimen which we made andfound the repetition stress amplitude ratio of eachtest specimen, the repetition loading number oftimes. Furthermore, we demanded RL20 andcompared the improvement effect of the liquefaction
strength. In addition, we performed aciddecomposition using HCl (0.5 mol/L) about the test
specimen after the examination. We consideredliquefaction strength and a CaCO3 separation rate,
the relations of the enzyme activity level.

By the calcium carbonate method which we paid
my attention to in this study, we produce hydrolysisof the urea by the metabolism of the microbe and let
you precipitate calcium carbonate all over the gap of
the sand. We write down a chemical reaction
formula in expression (1),(2). In hydrolysis inexpression (1), an enzyme activity level of the
microbe becomes important.
CO(NH2)2+2H2O→2NH4++CO32-(1)

Ca2++CO32- →CaCO3 ↓(2)

We carried out an examination about four kinds of
the Toyoura sand which saturated in pure water, the
Toyoura sand which we solidified by an isolated
microbe from all over the soil in Abashiri city,
Hokkaido, the Toyoura sand which we solidified by

Table 1 List of examination cases
CASE A
Microbe of the Abashiri
city, Hokkaido origin

Bacterial strain
Sand class
Relative density
Test specimen height
Test specimen diameter
The passing solidified
liquid number of times
Passing water interval

CASE B
Microbe of the Imizu city,
Toyama origin
Toyoura sand
Dr=50%
100mm
50mm

CASE C
Bacillus pasteurii

Three times
72 hours

Table 2 Composition of the solidified liquid
Reagent name
Nutrient BrothNH4Cl
NaHCO3CO(NH2)2
CaCl2
Pure water

Weight
0.3g
1.0g
0.21g
1.80g
3.33g
100ml

Fig. 1 Point that isolated a microbe

Table 3 An enzyme activity level of the microorganisms

Enzyme activity level(U/L)

Microbe of the Abashiri
city, Hokkaido origin
28
597

Microbe of the Imizu
city, Toyama origin
74

Bacillus pasteurii
654

GEOMATE- Bangkok, Nov. 14-16, 2016

RESULTS40
Deviator stress q(KPa)(1) Comparison of liquefaction properties
We write down the stress distortion relations thatthe loading number of times compared about a casebecoming equal repeatedly of test specimen and
CASE A,B,C which saturated only Toyoura sand inpure water in figure -2. Because a tendency that axis
distortions gradually increase together and liquefy
CASE A,B,C from a figure is recognized, we am
guessed when a solidification effect by the CaCO3
separation promotion derived from a microbe isprovided. Then, we write down the repetition
loading number of times in DA=5% in eachexamination CASE and the relations of the shear
stress ratio in figure -3. It followed that we could
expect a liquefaction strength improvement effectmost about CASE C using Bacillus pasteurii where
the effectiveness became clear than a study of thepast from a figure. In CASE A,B, an improvement
effect of the liquefaction strength was accepted
equally, but less than CASE C it followed. It isthought that an enzyme activity level of the microbe
which we used for CASE A,B as a cause is low.

Imizu city

30
20
10
0
-10
-20
-30
-40
-10

-5

0

5

10

Axial strain εa(%)

CASE B

Deviator stress q(KPa)40(2) Comparison of the CaCO3 separation rate
We write down the value of the CaCO3
precipitation rate for the sand weight of each test
specimen which we got from acid decomposition intable -4. It was intended to clarify the influence thatthe difference in microbe class gave for crystal
separation. But it was revealed that a CaCO3
separation rate did not have a big difference by amicrobe class as we showed it in table -4. Influenceby the solidification situation (including theunevenness in crystal size and the test specimen of
calcium carbonate) is thought about than the above.

Bacillus
pasteurii

30
20
10
0
-10
-20
-30
-40
-10

-5

0

5

10

5

10

Axial strain εa(%)

CASE C40
Deviator stress q(KPa)

Deviator stress q(KPa)

40
Abashiri city

30
20
10
0
-10
-20

30
20
10
0
-10
-20
-30
-40

-30

-10

-40
-10

-5

0

5

-5

0

Axial strain εa(%)

10

Axial strain εa(%)

Toyoura sand

CASE A

Fig. 2 The stress distortion relations that theloading number of times
598

GEOMATE- Bangkok, Nov. 14-16, 2016

DISCUSSION

1) The difference occurs to liquefaction strength byurease activity value.

From this test result, an improvement effect of
the liquefaction strength due to the isolated microbe
became clear from all over the soil in Abashiri city,
Hokkaido and Imizu city, Toyama. The
improvement of the liquefaction strength due to theoriginal position microbe is guessed regardless of apoint when at the same level. About both points thevalue of RL20 less than B.pasteurii it followed. We
paid my attention to a CaCO3 separation rate as acause, but the value of the acidolysis became at thesame level. Based upon the foregoing, it is guessedthat an enzyme activity level of the microbe to use
for microbe solidification affects the separation
place of CaCO3 in the test specimen. Furthermore, it
is different in a trend after having injected it in a test
specimen, and the possibility that a difference andthe deflection of the movement place affect the
CaCO3 separation place is thought about by amicrobe class. Based on the above-mentioned
situation, we perform similar examination using the
microbe varying in the enzyme activity level, and itis thought that it is necessary to make an enzyme
activity level and relations of the liquefaction
strength clear more.

2) The improvement effect of the liquefaction
strength was admitted in Microbial CaCO3
precipitation using in-situ microorganisms in the
spots selected by this research.
In addition, we divide a test specimen into theupper part, the central part, the lower part andperform acid decomposition about each plans
pushing forward further examination about theinfusion methods to get a homogeneous
solidification effect.
REFERENCES[1] Mastromei, G., Marvasi, M., and Perito, B.:
Studies on bacterial calcium carbonate
precipitation for stone conservation, BGCE,
pp.103-105, Delft, 2008.
[2] Whiffin, V.S., Van Paassen, A.L. and Harkes,
M.P.: Microbial Carbonate Precipitation as a
Soil Improvement Technique, Geomicrobiology
Journal, 24, pp.417-423,2007.

Table 4 Value of the CaCO3 separation rate for thesand weight

CONCLUSIONS
We understood two things from examination
results.

CaCO₃
separationrate(%)

CASE A

CASE BCASE C

2.044

2.236

2.038

Cyclic stress ratio R=σd/2σ'0

0.6
0.5
0.4

CASE A

0.3

CASE B

0.2

CASE C

0.1

Toyoura sand

0
1

10

100

1000

Number of cycles Nc(cycles)

Fig. 3 Repetition loading number of times in DA=5%599

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

A STUDY OF RESTRAINT TECHNIQUES FOR CEMENT TREATED
SOIL’S DETERIORATION BY MICROBIAL FUNCTIONS
1,2

Kazuki Mihara1, Toshiro Hata2
Department of Environmental Engineering, Toyama Prefectural University, Japan
ABSTRACT

Cement treated soil is used as a countermeasure for poor ground. But, cement treated soil deteriorate bycalcium leaching because of exposing seawater. This study considered a technology of reducing deterioration of
cement treated soil by using urease-producing bacteria. Authors tested seawater exposure tests using 2 types of
the urease-producing bacteriums of Sporosarcina aquimarina isolated from the sea off the coast of Korea andBacillius pasteurii isolated from land. The purpose of this tests are checking effect of reducing deterioration of
cement treated soil by the urease-producing bacteria. At the same time, we performed tests to confirm the growthof the bacteria in cement treated soil. We tested seawater exposure tests 2 types of conditions of temperature
20℃ is used for exposure tests, and 30℃ is a temperature that is suitable for the growth of bacteria, the purposeof this tests are checking relation temperature and deterioration speed. The main outcomes are as follows:
1) 2 types of the urease-producing bacteria have an effect of reducing deterioration 2) 2 types of condition oftemperature have relation to deterioration speed.
Keywords: cement treated soil, reducing deterioration, seawater exposure test, Micro Carbonate Precipitation
combination to show in Table.2(case C’s liquidmedium include B.pasteurii, Case D’s liquidmediums include S.aquimarina), and tested it
according to a test flow to show in Fig.1. This test
pieces were cured in the air for 28 days under 20℃.
Test pieces were sealed the side and the base likeFig.2, only the top surface of test pieces exposure
seawater.

INTRODUCTION
In Japan, an alluvial bed develops in coastal zone,
there are poor grounds. Therefore, Ground
improvement is necessary to build the structure.
Today, cement treated soil is used as acountermeasure for poor ground. But, cement treatedsoil is deteriorate by the exposure of seawater.
Cement treated soil exposures to seawater,
deteriorate by calcium leaching because of exposing
seawater [1]-[3]. Therefore it is necessary to
examine long-term stability. This study considered atechnology of reducing deterioration of cement
treated soil by using urease-producing bacteria. Urea
is hydrolyzed by urease, generates ammonia,
carbonate ion, and hydrogen ion. This carbonate ion
reacts with leaching calcium to form CaCO3 in
cement treated soil. Eq. (1) is hydrolyzing urea byurase-producing bacterium. Eq. (2) is formation of
CaCO3’s precipitation mechanisms [4].
2(1)CO( NH2 )2 + 2H2O ® 2NH3 + CO3 + 2H +2-

Ca2+ + CO3 ® CaCO3 ¯

Table 1 Characteristics of bacteria
B.pasteurii

S.aquimarina

isolate source
salinity tolerance(NaCl %)
optimum pH

land

coast

10

13

9

anaerobiosis

+

6.5～7
+

Table 2 combination table

(2)

Case

A

B

pure water (%)

21.05

20.99

MATERIAL AND METHOD

blast-furnace slag
cement type B (%)

6.58

6.56

In this study, authors examed seawater exposure
test of cement treated soil. Authors tested seawater
exposure tests using 2 types of the urease-producing
bacteria of Sporosarcina aquimarina isolated fromthe sea off the coast of Korea and Bacillius pasteurii
isolated from land. Table.1 shows Both bacteria’s
characteristics [5]. Authors made test pieces bypotter's clay (%)

6.58

6.56

silica sand (%)

65.79

65.57

Urea (%)

0.00

liquid medium (%)

0.00

600

C

D

14.43

0.33
0

6.56

6.56

Ca value and Mg value (g)GEOMATE- Bangkok, Nov. 14-16, 2016

0.15
0.1
0.05
0
-0.05
-0.1
-0.15
-0.2
-0.25
-0.3
-0.35
0

Fig.1

30
60
Time (day)Test flow

90(a) Case A
Ca ialue and Mg value (g)Fig. 2

0.15
0.1
0.05
0
-0.05
-0.1
-0.15
-0.2
-0.25
-0.3
-0.35

seawater exposure test setup

0

Table 3 Seawater’s composition table

Ca2+ concentration (ppm)
Mg2+ concentration (ppm)pH90(b) Case B
0.15
0.1
0.05
0
-0.05
-0.1
-0.15
-0.2
-0.25
-0.3
-0.35

Ca value and Mg value (g)Including elements

30
60Time (day)Cl, N, C, Na, Ca,
Sr, Mg, K, B, S,Br
125
830
8.32

Seawaterused
artificial
seawaterofcommercially available. Table.3 shows Seawater’s
composition table. Test period is 84 days, seawater
was sampled every 14day, changing water andtesting needle penetration test every 28 days.
In needle penetration test, authors pricks the top
surface and bottom surface with a needle, measure 2
types of penetration value. 2 types of penetration
values difference assumed deterioration depth.
Sampled water was analyzed Ca ion and Mg ion by
Atomic Analyzer.
Authors tested seawater exposure tests 2 types ofconditions of temperature 20℃ is used for exposure
tests, and 30℃ is a temperature that is suitable forthe growth of bacteria, the purpose of this tests arechecking relation temperature and deterioration
speed.
Therefore, in case D, authors measured urease
activity when finished air curing 28days forconfirming the growth of bacteria under high alkali
condition in cement created soil. Put part of test
pieces in NH4-YE mediums (ATCC medium: 1376),
grow for a week, measured urease activity.

0

30
60
Time (day)90(c) Case C
Ca value and Mg value (g)

0.15
0.1
0.05
0
-0.05
-0.1
-0.15
-0.2
-0.25
-0.3
-0.35
0

30

60

90

Time (day)(d) Case D
0.15

30℃-Ca value(g)020℃-Mg-0.35 value(g)50
30℃-Mgvalue(g)カル
シウ
ム…

20℃-Ca value(g)

Fig.3
601

Ca value and Mg value

GEOMATE- Bangkok, Nov. 14-16, 2016

Deterioration depth (mm)RESULTS AND DISCUSSIONSCa leaching and Mg absorption
Fig.3 shows Analyzing results of Ca value and Mg
value. This figure shows that it is leaching from test
pieces when plots are increasing, and it is absorption
from test pieces when plots are degreasing. In case
A to D, Ca’s plots are increasing, so Ca is leachingfrom test pieces, Mg’s plots are decreasing, so Mg isabsorption to test pieces. Moreover, 30 ℃ of
temperature condition’s Ca leaching and Mg
absorption is more than 20 ℃ of temperature
condition. From those trends, it is thought thattemperature condition has relation to deterioration.
Therefore, in case C and D of including ureaseproducing bacteria, Ca leaching and Mg absorption
value restrain in compared with case A and B thatcase without urease-producing bacteria.

Deterioration depth (mm)Deterioration depth (mm)

0

60

90

0.00
0.50
1.00
1.50
2.00
2.50
3.00
3.50
4.00
4.50
5.00

30

60

90

60

90

Time (day)(c) Case C

Deterioration depth (mm)0Table 4 Urease activity of case D(including S.aquimarina)

0.00
0.50
1.00
1.50
2.00
2.50
3.00
3.50
4.00
4.50
5.00

30

Time (day)(d) Case D-0.35
0.15 20℃
0

カ
ル

22.02

30

0.00
0.50
1.00
1.50
2.00
2.50
3.00
3.50
4.00
4.50
5.00

(b) Case B

Table.4 shows urease activity of case D. Urease
activity of after air curing is 22.02 (U/L). Urease
activity of after air curing decreased a one-fifth ofurease activity of liquid mediums before add to test
pieces, authors could be confirmed growth ofbacteria.

110.0890

Time (day)Growth of bacteria in cement treated soil

Ureaseactivity(U/L)

60

time (day)0

Fig.4 shows deterioration depth of case A~D. incase A~D, 30 ℃ of temperature condition’s
deterioration depth is more than 20℃ of temperature
condition. From those trends, it is also thought thattemperature condition has relation to deterioration.
Therefore, in case C and D of including ureaseproducing bacteria, deterioration depth restrain incompared with case A and B that case without
urease-producing bacteria. The factor affecting is
CaCO3 precipitated in test pieces’ gap.
Especially, in both temperature conditions, case C
is effective most in those cases.

After aircuring28days30(a) Case A

Deterioration depth

Before add
to testpieces

0
0.00
0.50
1.00
1.50
2.00
2.50
3.00
3.50
4.00
4.50
5.00

Fig.4
602

Deterioration value

30℃

50

GEOMATE- Bangkok, Nov. 14-16, 20162) 2 types of condition of temperature have relation
to deterioration speed. Deterioration speed becomes
higher with temperature condition become higher.
3) Urease-producing bacteria growth under highalkali condition in cement treated soil, ureaseproducing bacteria may show restraint deterioration
effects for a long term.
4) Analyzing calcium leaching value can estimate
deterioration value in non-destruction.

0.1

Ca value (g)

0.08
0.06
0.04
y = 0.0146x
R² = 0.7003

0.02
0
0.00

Fig.5

1.00 2.00 3.00 4.00
Deterioration value (mm)5.00

ACKNOWLEDGEMENTS
I would like to show my greatest appreciation to
Tomomi Takamatsu who offered continuing support
and constant encouragement. I would also like tothank Takumi Nagai and Kenta Teranishi whose
meticulous comments were an enormous help to me.
Finally, I would like to express my gratitude toGrants-in-Aid for Scientific Research foundation(B) No.26289154 for their financial support.

Relationship between Ca value anddeterioration.

Relation to Ca leaching value and deterioration
depth
Fig.5 shows a relationship between Ca leachingvalue and deterioration. As shown in Figs.5, there
are scattering, deterioration value increases withincrease in calcium leaching value. Ca leachingvalue related to deterioration value. From this,authors thought to be able to estimate deterioration
value by analyzing calcium leaching value in nondestruction.

REFERENCES[1] Hara S, Suestugu D, Hayashi S, “Calcium
Leaching Mechanism of Lime Treated Soil
Immersed in Seawater”, J. of the Society ofMaterials Science, Japan, Vol. 61, No.1, Jan.
2012, pp. 11-14.
[2] Nakamura T, Kitazume M, “Laboratory Tests
on Durability of Cement Treated Soils”,
TECHNICAL NOTE OF THE PORT AND
AIRPORT RESEARCH INSTITUTE, No.1128,
Jun. 2006.
[3] Hara H, Suetsugu D, Hayashi S, Matsuda H,“DETERIORATION MECHANISM OF
CEMENT-TREATED SOIL UNDERSEAWATER”, J of JSCE division C:
Geotechnics, vol.69, No.4, Apr. 2013, pp. 469479.
[4] Qabany, A.A, Soga, K., Santamaria, J.C.,“Factors Affecting Efficiency of Microbially
Induced Calcite Precipitation”, Journal of
Geotechnical and Geoenvironmental
Engineering, Vol.138, No.8, 2012, pp. 992-1001.
[5] Hata T, Yokoyama T, and Abe H,“Investigation of a soil improvement techniquefor coastal regions on the basis of calcium
carbonate precipitation by using the ureahydrolysis rate” J of Geotechnical engineering,
Vol.8, No.4, Jan, 2013, pp. 505-514.

CONCLUSIONSIn this study, Authors considered a technology of
reducing deterioration of cement treated soil byusing urease-producing bacteria. Authors tested
seawater exposure tests using 2 types of the ureaseproducing bacteria of S.aquimarina and B.pasteurii.
At the same time, we performed tests to confirm thegrowth of the bacteria in cement treated soil. We
tested seawater exposure tests 2 types of conditionsof temperature 20℃ is used for exposure tests, and
30℃ is a temperature that is suitable for the growthof bacteria. In addition, in case D, authors measured
urease activity when finished air curing 28days forconfirming the growth of bacteria under high alkali
condition in cement created soil.
The main outcomes are as follows:
1) 2 types of the urease-producing bacteria have aneffect of reducing deterioration. Especially, case C iseffective most in those cases, B.pasteurii can be
expected restraint techniques for cement treatedsoil’s deterioration.

603

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

REMOVING FLUORIDE FROM A HOT SPRING USING AN
ELECTROLYSIS SYSTEM
Yuki Imai, Shiori Yanagawa, Misa Konishi, and Tomonori Kawakami
Faculty of Engineering, Toyama Prefectural University, Japan
ABSTRACT
A high concentration of fluoride in the wastewater from hot springs is an environmental issue in Japan, sincesome of the wastewater exceeds the national minimum effluent standards of 8 mg/l. However, an effective
treatment for fluoride removal has not yet been developed. Accordingly, the temporal effluent standards of
15mg/l - 50 mg/l have, so far, been applied to the wastewater from hot springs.
In this study, an electrolysis system consisting of an anode bath and a cathode bath separated by a diaphragm
made of a clay panel was tested for the removal of fluoride. In an electrolysis system, fluoride is removed by coprecipitation with magnesium hydroxide formed in a cathode bath under a high pH condition.
As a pretreatment of wastewater, 100 mg/l of magnesium was added to water from Gero hot spring, Gifu,
Japan, to enhance the formation of the precipitation of magnesium hydroxide, since water from Gero hot springcontains less than 1 mg/l of magnesium. Water from Gero hot spring to which 100 mg/l of magnesium had been
added was treated by an electrolysis system with a flow rate of 10 l/day and a current of 120 mA. The
electrolysis system reduced the fluoride concentration from 16.6 mg/l to 6.4 mg/l, which meets the national
minimum effluent standards of 8 mg/l.
Keywords: Gero hot spring, co-precipitation, sequential flow reactor, national minimum effluent standard

appropriate technologies for effectively removing
fluoride [3]. There are some techniques to remove
fluoride from drinking water such as reverse osmosis[4], electrodialysis and nanofiltration [5-6], however,these are the techniques just separate fluoride into
the solution with higher fluoride concentration andthe solution with lower concentration. These
techniques are not suitable for wastewater treatment.
To remove fluoride from industrial wastewater, theprimary treatment has generally been to add calcium
to produce insoluble CaF 2 , followed by coagulation.
According to the solubility product of CaF 2 (3.9×10-11 mol3/l3), 8.8mg/l of calcium is sufficient toobtain 8 mg/l of fluoride theoretically; however, in
the actual cases, an 8-mg/l concentration of fluoride
cannot be achieved even by adding 300 mg/l ofexcess calcium. As a result, secondary treatments,
such as ion exchange or the addition of aluminum,
are introduced to meet the regulation [7-9].
Gero hot spring is located 35°48’N and 137°14’E
in Gifu Prefecture, Japan. It is widely regarded as
one of the three best hot springs in Japan [10]. Morethan 40 hotels offer rooms and hot spring baths. Hot
water is pumped out to deliver to the hotels.
Temporary effluent standards of 30 mg/l and 15
mg/l are applied to small and large hotels,
respectively. The fluoride concentration, which is
officially reported to be 16.5 mg/l [11], exceeds
regulations for large hotels. However, the abovementioned treatments to remove fluoride are quiteprohibitive for the hotel business due to the high costcompared to their business scale.

INTRODUCTION
UNICEF has designated Japan as one of thecountries with endemic fluorosis due to excess
fluoride in drinking water [1]. The health problem
has been remedied by using alternative water
sources that are free of fluoride. However, highconcentrations of fluoride in the wastewater remain
an environmental issue in Japan, since somewastewaters exceed the national minimum effluent
standards of 8 mg/l [2]. Since no effective treatment
for the removal of fluoride has been developed,
temporal effluent standards have been applied to thewastewater of some industries. The operation of
hotels with hot springs, very popular with Japanesepeople for vacations, is one such industry, since thewastewater discharged by the hotels sometimes
contains a high concentration of fluoride originatingfrom the hot spring. Different temporal effluent
standards for fluoride concentration are applied tohotels with hot springs depending on the amount of
the effluent and the type of hot spring. Hotels thathave an effluent of more than 50 m3/day have a
temporal effluent standard of 15 mg/l. For the otherhotels, different standards, such as 30 mg/l or 50
mg/l, are applied based on the type of hot spring, i.e.,
pumped out or gushed out naturally, respectively.
The possibility of stiffening temporal effluent
standards has been discussed by the Japanese
Ministry of Environment every three years; however,as decided at the council held in 2015, no changes
will be made in the regulation due to the lack of
604

GEOMATE- Bangkok, Nov. 14-16, 2016

In the current study, a new electrolysis system was
examined as a cost-effective system for removing
fluoride from wastewater from hotels with hot
springs to the level of the national minimum effluent
standard of 8 mg/l. We focused on the water of Gero
hot spring as a model.

Alkalinity was removed from the solution by adding
sulfuric acid in accordance with the following
equation:HCO 3 - + H+ → CO 2 + H 2 O.
CO 2 formed and dissolved in the solution wasremoved by aeration.
Using the solution mentioned above, the first of a
two-series electrolysis was performed. After the firstelectrolysis, part of fluoride was co-precipitated withmagnesium hydroxide in the cathode bath. At thesame time, part of the fluoride was transferred to theanode bath by the Coulomb force. Since one- andtwo-sided treatments are both required for
wastewater treatment, a second electrolysis was
performed to treat the solution in the anode bath.
The solution taken out of the anode bath was mixed
with the same amount of the raw hot spring water towhich magnesium had been added. The second
electrolysis was performed with this solution. The
addition of sulfuric acid was not needed because thesolution taken out of the anode bath was sufficiently
acidic to remove alkalinity.

MATERIALS AND METHODS
Gero hot spring
Raw hot spring water from Gero was used for
fluoride removal. The objective of the current
research is the removal of fluoride from wastewater;
however, raw hot spring water was used for theexperiments, since it has the highest concentration offluoride without dilution. Fluoride removal was
examined in a laboratory using a batch reactor and asequential flow reactor.
Batch reactor
Electrolysis using a batch reactor was performed
to remove fluoride. The batch reactor shown in Fig.
1 consisted of two electrolysis baths separated by amembrane filter into a cathode bath and an anodebath. Each bath had a volume of 300 ml. In anelectrolysis system, fluoride is removed by coprecipitation with magnesium hydroxide formed in
the cathode bath under a high pH condition.
Magnesium was added to the water from Gero hot
spring to enhance the formation of magnesiumhydroxide, since the magnesium concentration ofwater from Gero hot spring, less than 1 mg/l, is too
low for the formation of magnesium hydroxide. The
magnesium concentration was adjusted to 50 mg/l
with magnesium chloride. A constant current power
supply, adjusted to 80 mA, was used for theelectrolysis. Electrolysis was performed for 1 hour.
Since it was found that alkalinity (HCO 3 -)interfered with the formation of magnesiumhydroxide by forming magnesium carbonate,alkalinity was removed before electrolysis.

Sequential flow reactor
As a sequential flow reactor, a reactor with ananode bath and a cathode bath separated by a clay
panel was used (Fig. 2). The volume of each bath
was 420 ml. An aeration bath was installed at the
inlet of the cathode, where alkalinity was removedby the acid solution formed in the anode bath. Waterfrom Gero hot spring to which magnesium had been
added in advance was introduced to the anode andaeration baths. The total flow rate was set at 10 l/day,which is equivalent to a retention time of 2 hours. In
the experiments, the flow ratio of Gero hot springwater into the aeration bath (b) to the anode bath (a)was varied with 4 ratios, i.e., b:a=0:10, 5:5, 8:2, and9.4:0.6. The Mg concentration was adjusted to 100
mg/L. One hundred twenty mA was applied forelectrolysis with a constant current power supply.

Fig. 2

Fig. 1

Schematic flow of the batch reactor

605

Schematic diagram of the sequential flow
reactorGEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 Water quality of Gero hot spring. The concentration for 2013 is from the official record
Date
Y/M/D

pH

EC
mS/m

Na
mg/l

NH4
mg/l

K
mg/l

Mg
mg/l

Ca
mg/l

F
mg/l

Cl
mg/l

NO3
mg/l

SO4
mg/l

Alkalinity
meq/l

2015/ 9/11
2013/11/ 5

8.9
9.5

35
52.2

107.2
108.9

0
0

1.0
1.2

0.1
0

0.6
1.9

17.7
16.5

75
75

0
0.1

10.8
10.9

1.66
1.65

Analysesmagnesium concentration decreased from 45.9 mg/l
to 19.7 mg/l and white precipitation appeared in thecathode bath, there could be little co-precipitation offluoride with magnesium hydroxide, due to theformation of magnesium carbonate.
The fluoride removed from the hot spring water
without alkalinity by the first electrolysis is shownin Table 3. Due to the addition of sulfuric acid toremove alkalinity, the pH value of the initial solution
was low. The fluoride concentration decreased in the

Major ions were measured by an ion
chromatography after samples were filtered by amembrane filter with a pore size of 0.45 μm. The pH
was measured by a glass electrode method. Titration
to the endpoint of pH=4.8 by sulfuric acid was used
to measure alkalinity.
RESULTS AND DISCUSSION
Gero hot spring

Table 2 Fluoride removed by electrolysis withalkalinity
pHF
Mg(mg/L)(mg/L)
Initial
8.8
16.9
45.9
3
Anode
2.4
22.3
37.7
3
Cathod
11.
7.0
1.7
e
1
Averag
14.7
19.7
e

The quality of water from Gero hot spring
sampled on September 11, 2015, is shown in Table 1,
together with the official record sampled on
November 5, 2013. The official record is theanalytical concentrations of chemical components
registered in Gifu Prefecture based on Japan’s Hot
Spring Law.
The concentrations of the major ions, including
alkalinity, of our analysis were quite similar to thoseof the official record, with the exceptions of calcium
and fluoride. The concentration of fluoride
fluctuated in the other samples between 15.8 mg/L
and 18 mg/L. In the sample from September 11,
2015, the fluoride concentration was as high as 17.7
mg/L, which is approximately twice the national
minimum effluent standard of 8 mg/L. The Mg wascontained slightly, the pH was 8.9, and the alkalinity
was 1.66 meq/L.

Table 3 Fluoride removed by the first electrolysis
without alkalinity
pHF
Mg(mg/L)(mg/L)
Initial
3.9
15.8
47,1
Anode
2.3
19.6
28.7
Cathod
11.
5.2
0.0
e
6
Averag12.4
14.4
eBatch reactor
Table 2 shows the fluoride removed byelectrolysis from the hot spring water with alkalinity.
The pH decreased in the anode bath while it
increased in the cathode bath. The fluoride
concentration decreased in the cathode bath;however, it increased in the anode bath, indicatingthat fluoride was transferred from the cathode bathto the anode bath by the Coulomb force.
Accordingly, the decrease in the fluoride
concentration was very small when the averaged
concentration of both baths (14.7 mg/l) after
electrolysis is compared with the initialconcentration (16.9 mg/l). Even when the average

Table 4 Fluoride removed byelectrolysis without alkalinity
pH
F(mg/L)
Initial
2.8
17.2
Anode
2.3
19.4
Cathod11.
8.3
e0
Averag
13.9
e
606

the

secondMg(mg/L)
49.2
42.3
0.4
21.4

GEOMATE- Bangkok, Nov. 14-16, 2016expressed in flow rates of the anode bath of 10, 5, 2,
and 0.6 l/day, respectively.
A lower flow ratio resulted in higher fluoride
removal. In the case of the ratio of 9.4:0.6, thefluoride removal rate was 62%, and the fluoride
concentration after treatment was 6.4 mg/l; as aresult, the national minimum effluent standard of 8
mg/l was achieved. The relationship between pH andfluoride concentration and the relationship between
pH and magnesium in the cathode bath are shown in
Fig. 4.
When pH was higher, the fluoride removal rateincreased and the magnesium concentration
decreased. The change in pH, the concentration offluoride, and the magnesium concentration in the hotspring water, the anode bath, the aeration bath, andthe cathode bath are shown in Figures 5, 6, and 7,
respectively, at flow ratios of b:a=0:10 and 9.4:0.6.

cathode bath but increased in the anode bath,indicating that fluoride was transferred in the samemanner as with alkalinity. However, some amount offluoride was found to have been removed by
precipitation during electrolysis, since the average
concentrations of fluoride in both baths after
treatment were well below that of the initial solution.
When the average concentrations of the anode andcathode baths are compared, electrolysis without
alkalinity was showed to more effectively remove
fluoride than electrolysis with alkalinity.
The result of the second electrolysis is shown in
Table 4. After treatment, the fluoride concentrationin the cathode bath decreased to 8.3 mg/l. The
fluoride concentration in the anode bath was 19.4
mg/l, which was similar to the concentration offluoride in the anode bath with the first electrolysis.
This indicates that the fluoride concentration could
be decreased to approximately 8 mg/l when batch
electrolysis is repeated.
Sequential flow reactor
According to the experimental results of thebatch reactor showing that repeated batch
electrolysis could remove fluoride, a sequential flow
reactor was tested for removing fluoride.
The effect of the flow ratio on the removal offluoride is shown in Fig. 3. In the figure, the flow
ratios of b:a=0:10, 5:5, 8:2, and 9.4:0.6 are

Fig. 5

Change in the pH along the flow

Fig. 6

Change in the fluoride concentration along
the flow

Fig. 7

Change in the Mg concentration along theFluoride removal ratio (%)

70
60
50
40
30
20
10
0
0

Fig. 32

4
6
8
Flow rate of anode bath (L/day)

Effect of the flow ratio on the removalof fluoride

F concentration(mg/l)

16

100

14

80

12
10

60

8
6

40

F

4

Mg

2

20

0

0
7

Fig. 4

8

9
pH

10

11

Relationship between pH and fluoride
concentration and the relationshipbetween pH and magnesium in thecathode bath607

GEOMATE- Bangkok, Nov. 14-16, 2016

flow.
When the flow ratio was 0:10, the fluoride
concentration in the treated water slightly decreased
from that in the hot spring water. When comparedwith the batch reactor, the pH was as low as 8.0 in
the cathode bath, and the concentration of dissolved
magnesium was high. The reason is that the pH did
not recover in the cathode bath after the pH in the
mixing bath decreased to 2.35. Therefore,
magnesium hydroxide did not form well, and coprecipitation with fluoride did not take place. Whenthe flow ratio was 9.4:0.6, acidic solution with a pH
value of 1.8 was formed in the anode bath. However,
due to smaller volume of the acid solution thatflowed into the aeration bath, a higher pH value of
2.60 was obtained in the aeration bath. As a result,
pH increased to 10.5 at the outlet of the cathode bath,and the fluoride concentration decreased to less than
the standard concentration of 8 mg/l. The
concentration of dissolved magnesium decreased to14.6 mg/l, as well. The difference in the pH valuesof 2.60 and 2.35 in the aeration bath between caseswith flow ratios of 9.4:0.6 and 0:10, respectively,
seems not to be significant. However, this is because
pH is expressed in a logarithmic scale. When pH
values are converted into hydrogen ion
concentrations, a pH of 2.35 is equivalent to 4.5mmol/l of H+, while a pH of 2.60 is equivalent to 2.5mmol/l of H+. The difference in the concentrationsof H+ as 2.0 mmol/l was large enough to cause a
difference in the OH- concentration in the cathodebath of 0.3 mmol/l, which was calculated from thedifference between pH=8.0 and pH=10.5.

added to the solution might be reused. When the hotspring water contains magnesium naturally, it is notnecessary to add magnesium; however, theprecipitation that does not need to be reused, in thiscase, will generate sludge. Nevertheless, the amountof sludge would be much less than that when a Caadding method is used, since excess Ca should be
added to the solution. In spite of this, reducing
electricity, magnesium usage, and sludge would bechallenges for the future.
CONCLUSION
The electrolysis system for treating Gero hot
spring water to remove fluoride was tested in alaboratory. The sequential flow reactor could
remove fluoride to a level of 6.4 mg/l, which meetsthe national minimum effluent standards of 8 mg/l. It
would be a cost-effective method for removing
fluoride from wastewater.
ACKNOWLEDGEMENTS
This research was partly supported by JSPS
KAKENHI Grant Number 15H05120 and Center ofCommunity (COC) program by MEXT.
REFERENCES[1] Qian, J., A. K. Susheela, A. Mudgal and G. Keast
G. F1luoride in water: An overview, Waterfront:
A UNICEF Publication on Water, Environment,
Sanitation and Hygiene, 1999, 13, 11-13[2] http://www.env.go.jp/water/onsenhaisui/conf/11.html[3] Japan’s Ministry of theEnvironment, http://www.env.go.jp/water/onsenhaisui/conf/12.html[4] Peter Sehn, Fluoride removal with extra low
energy reverse osmosis membranes: three yearsof large scale field experience in Finland,
Desalination, Vol. 223, Issues 1–3, 73-84, 2008
[5] Zakia Amor, Suad Malki, Mohamed Taky,
Bernard Bariou, nabil Mameri, Azzeddine
Elmidaoui, Optimization of fluoride removalfrom brackish water by electrodialysis,
Desalination,120, 263-271, 1998
[6] Simons, R., Trace element removal from ash
dam waters by nanofiltration and diffusion
dialysis, Desalination, Vol. 89, Issue 3, 325-341,
1993[7] Yuji Wada, The high class treatment of fluoridecontaining wastewater, Kagaku To Kogyo 76(11),557-564, 2002
[8] Paripurnanda Loganathan, Saravanamuthu
Vigneswaran, Jaya Kandasamy,Ravi Naidu,
Defluoridation of drinking water using
adsorption processes, Journal of Hazardous

Cost and maintenance
We proved that with this electrolysis system, it ispossible to remove fluoride from hot spring
wastewater. In the proposed electrolysis system, thedevices can be made inexpensively, since thediaphragm of the electrolysis baths is made of a clay
panel that would be robust against fouling, as well.
Electric consumption for the electrolysis wasapproximately 1 W, excluding consumption by thewater pumps, indicating that 24 Wh would be
required for the treatment of 10 L of Gero hot springwater in one day. Supposing that 1 kWh=17
Japanese yen, the treatment of 10 L would require
only 0.4 Japanese yen.
It could be possible to reuse magnesium bylowering the pH to dissolve the precipitation. The
recovery ratio of magnesium as precipitation was ashigh as 85%, indicating that most of the magnesium608GEOMATE- Bangkok, Nov. 14-16, 2016

Materials, 248-249, 1-19, 2013[9] Sachin V. Jadhav, Eugenio Bringas, Ganapati D.

[10] Japan National Tourism Organization, JAPAN:
The Official Guide,

Yadav, Virendra K. Rathod, Inmaculada Ortiz,
Kumudini V. Marathe, Arsenic and fluoride
contaminated groundwaters: A review of current
technologies for contaminants removal, Journalof Environmental Management, Vol. 162, 1, 306325, 2015

http://www.jnto.go.jp/eng/location/regional/
gifu/gero.html[11] Gifu public health inspection center, Analyticalreport of hot spring, Gero hot spring (No.125),
2013

609

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

GENETIC DIVERSITY AND GENETIC STRUCTURE OF AN
ENDANGERED SPECIES, ERIOCAULON NUDICUSPE, GROWING IN
ARTIFICAL DISTURBING HABITATS
1,2Michiko Masuda1, Tadamasa Fukagawa2 and Fumitake Nishimura3
Department of Civil Engineering, Nagoya Institute of Technology, Japan; 3Department of Environment
Engineering, Kyoto University, Japan

ABSTRACT
Eriocaulon nudicuspe (Eriocauloaceae) is an endangered species in Japan. Habitats of the species are lost bycity development. So many conservation areas are established in Aichi Prefecture, and protection activities aredone by many natural protecting groups, and the extinction rate in recent years is being decelerated. But many
protecting groups often transplanted from some other places, and genetic disturbance was a problem. We
sampled 12 populations of the species from nature reserved areas, and studied them for allelic variation at 17
enzyme loci. There was no significant correlation between the real distance of conserved areas and genetic
distances, suggesting that the gene disturbances occurred in these areas especially in frequently managed areas.
On the other hand, the degree of the genetic differentiation at strictly conserved area where conservation
management is done only once a year was high and there was no evidence about genetic disturbances. There wasa possibility peculiar genetic disappearance of each habitat for genetic disturbance, and the necessity with whicha guideline of protection activity is made was indicated.
Keywords: Genetic disturbance, Genetic distance, Conservation management, Nature reserved area,
Endangered species

where the genetic boundaries were shown in maps.
But there is no herbaceous plants guideline abouttransplantation because of the difficulties of thesamplings. Then we investigated the genetic
structure and genetic diversity of Eriocaulon
nudicuspe and analyzed the effect of theconservation activities with 3 questions, 1) Is there arelationship between the location and genetic
distance? 2) Is there a relationship between theconservation activities and genetic diversity? 3) Isthere a relationship between population size andgenetic diversity? It was paid attention to the threepoints and analyzed.

INTRODUCTION
Ueda (1989) defined Tokai hilly land element, 14
species plants, a lot of local endemic, semi-endemic
and relict taxa growing in small wetlands in theCentral Japan [1]. The habitats of the elements havebeen destructed by the urban growth, and all species
are defined endangered species. Tomita (2004)
reported the distribution change of Eriocaulon
nudicuspe for 25 years, almost 50% areas are
disappeared [2]. So Aichi Prefecture supported theconservation activities of the species.
For the conservation, the moderate disturbance isvery important, so many nature protecting groups
keep being active for decades [3][4]. The extinction
rate of the endangered species is less than 5 % for
the activities, and another population was appeared.
Everyone easily collects seeds of the species andgrows up by sowing. Sometimes people can easily
move seeds of the species attaching the sole.
But there is another problems occurred in the
disturbance. Some researchers pointed out thegenetic disturbance affect serious damage of theendangered species [5][6]. The conservation
activities sometimes transplant the endangered
species from one habitat to another habitat. This
manipulation causes hybrid breakdown, dilution andoutbreeding depression [7]-[9].
Tsumura and Suyama (2015) published the book
“The guideline for transplanting of trees seedling”

MATERIALS AND METHODSThe Study Site
The study was carried out on the wetland at Aichi
Pref. and Shizuoka Pref. (Table 1). 36 habitats of thespecies reported in the previous research [2], but
growing could be confirmed only in 17 habitats in2007. The investigation was done at 12 populations
that can secure the enough number of individuals.
Fig.1 shows the locations of 12 populations thatwere investigated. The capital letter of each location
exhibits same river systems.

610

GEOMATE- Bangkok, Nov. 14-16, 2016

plants listed the species a ‘vulnerable’ level species[10][11]. This species is distributed in acid marshwith low nutrient, wetland around paddy fields andedge of reservoirs around Ise-Bay area. The seeds ofthe species emerge in spring and bloom late summer
to early autumn. An individual has 1-10
inflorescence and produces 0-12 seeds per
inflorescence.

Table 1 The characters of study site, location
Studylocation
site
North
East
A1
34o 44’ 45”
137o 27’ 09”
A2
34o 42’ 10”
137o 24’ 58”oA3
34 38’ 38”
137o 16’ 30”oB1
34 55’ 04”
137o 46’ 20”
B2
34o 49’ 38”
137o 45’ 54”oB3
34 48’ 56”
137o 45’ 56”oC1
34 56’ 06”
136o 56’ 14”oC2
34 55’ 11”
136o 53’ 00”
C3
34o 54’ 00”
136o 52’ 15”oC4
34 52’ 07”
136o 53’ 53”oC5
34 50’ 09”
136o 53’ 10”oD1
35 08’ 56”
137o 04’ 57”
D2
35o 06’ 10”
137o 00’ 05”oE1
35 14’ 43”
137o 03’ 08”oE2
35 14’ 08”
137o 02’ 35”oF1
35 06’ 01”
137o 14’ 23”
F2
35o 07’ 52”
137o 09’ 30”
The same capital indicates the same water system

Fig. 2

The flowers of E. nudicuspe.

Electrophoresis
Fresh leaves were collected from 30 individuals
per population at June in 2007. Leaves were kept on
ice during 2 hours transportation to the laboratory.
The following enzyme systems were examined:
aconitase (ACO), sikimate dehydrogenase (SKDH),
iso-citrate
dehydrogenase(IDH),
malatedehydrogenase (MDH), acid phosphate (ACP),
phospho-glucose isomerase (PGI), phosphoglucomutase (PGM) and menadione reductase (MR).
Leaves were used to resolve the following 17
putative loci: aco-1, aco-2, skdh, idh, mdh-1, mdh-2,
mdh-3, acp-1, acp-2, acp-3, acp-4, acp-5, pgi-1, pgi2, pgh-3, pgm and mr. Samples were ground in acold extraction buffer described by Odrzykoski and
Gottlieb [12]. The enzymes were resolved on 10.8%
starch gel. System 5 of Soltis et al. [13] were used.
Staining procedures followed previous works [13][15].

E2。。E1
F2。
。。
。
D2D1 F1
C1。
C4。
。
C5

A2。

B3。
。
B2

Fig. 1. Distribution of the population examined.

The Statistical analysis
For each population the number of alleles per
locus (A), proportion of polymorphic loci (P), andgene diversity (h) were calculated. We used all locidata in the calculation of A, and regarded a locus aspolymorphic if the frequency of its most frequent
alleles is under 0.95. In addition, total gene diversity[16] was calculated for species level. The population

The Study Plant
Eriocaulon nudicuspe Maxim. (Fig. 2) is anannual plant distributed in circum-Ise Bay area, thecenter of Japan [1]. However its population has been
decreasing and Red Data Book of Japanese vascular611

GEOMATE- Bangkok, Nov. 14-16, 2016

And total gene diversity (H T ) of the species was
0.293. The levels of genetic diversity in E.
nudicuspe was almost same that of other endangered
species, for example, Aster kantoensis growing in
the river bed were 0.36(P), 1.53 (A) and 0.142 (h)[18]. And other endangered species showed, P(0.199 to 0.65), A (1.44 to 2.01) and h (0.037 to0.43) [19]-[22]. But the index of river wetland
endangered species, Penthorum chinense, is higher
than that of E. nudicuspe [23], these were 2.42 (A),
0.75(P) and 0.308 (h).

genetic structure was analyzed by initiallycalculating Nei’s G ST value [16]. Values for genetic
identities (I) and standard genetic distance (D) werecomputed for each pairwise comparison of all
populations. The neighbor joining method [17]based on D was used for constructing a phenogramfor E. nudicuspe.
RESULTSThe Study Site

Table 3. Mean number of polymorphic loci (A),proportion of polymorphic loci (P), and genediversity within a population (h) at 17 loci for
examined populations of E. nudicuspe

The number of individuals and the frequency ofconservation activities were shown in Table 2. The
number of individuals ranged from 20 to 50000. C5
and F2 wetland are left, and few persons hardly
enters. C4 and F1 are managed severely, andprotection activity is permitted only once a year.
And in the other area conservation activity is
performed, and reaping and organic matter removal
are conducted periodically.

Population
A2
B2
B3
C1
C4
C5
D1
D2
E1
E2
F1
F2

Table 2. The number of individuals of each
population in 2007, Geographical Features, andfrequency of conservation activities within a year
Study
No. ofGeograhical
Freq.
site
individuals
Featuresof act.
4
A1
1000
silt, clay, gravel
10
A2
1000
Terrace sediment
4
A3200
Terrace sediment
12
B1
50000
silt, clay, gravel
4
B2
1500
silt, clay, gravel
4
B3
30000
silt, clay, gravel
C1
1430
silt, clay, gravel
1
C2
20
mud marsh
1
C3
800
4
silt, clay, gravel
C4
5190
1
silt, clay, gravel
C5
3000
0
silt, clay, gravel
D1
7200
12
kaoline, gravel
D2
4000
12
graniteE1
1000
1silt, clay, gravel
E2
5000
1
silt, clay, gravel
F1
7000
1
silt, clay, gravel
F2
5200
silt, clay, gravel
0

P

A

0.647
0.765
0.706
0.800
0.818
0.765
0.294
0.588
0.412
0.588
0.706
0.471

2.059
2.118
1.941
2.400
2.636
1.941
1.471
1.941
1.529
1.765
2.059
1.529

h
0.156
0.165
0.129
0.350
0.473
0.141
0.078
0.194
0.078
0.116
0.157
0.079

Genetic diversity

3
p
A

2

h
1
0
0

10000 20000 30000
Population size

40000

Fig. 3. The relationships between population sizeand genetic diversity indexes p, A and hFig. 3 showed the relationships between
population size and genetic diversity. There are no
significant relationships between the population sizeand genetic diversity.
Fig. 4 showed the genetic distance among
population. The genetic distance between the nearest
populations were small, but the longest distantpopulation. Especially B2, B3 are the edge of the
distribution of the species, but they close to thecenter of the distribution. It is suggested that geneflow occurred constantly among the populations.

Genetic Diversity
Seventeen loci were scored: aco-1, aco-2, skdh,
idh, mdh-1, mdh-2, mdh-3, acp-1, acp-2, acp-3, acp4, acp-5, pgi-1, pgi-2, pgh-3, pgm and mr, fifteenloci were polymorphic. In all population aco-1 and
mr were monomorphic. Allele frequencies at thepolymorphic loci are listed in Appendix.
Table 3 summarizes the resultant values of A, P
and h for each population.
612

GEOMATE- Bangkok, Nov. 14-16, 2016

The result of total population genetic structure
G ST was 0.236. The value of G ST suggested that
middle level differentiation occurred. It is always
observed isolated populations (no differentiation: 00.05, low level differentiation: 0.05-0.15, middle
level: 0.15-0.25, high level: 0.25-).

of populations, which indicates that the effective
population size is independent of the actual
population size. Several factors are known to reduce
the effective size of population [24]: (1) fluctuationin population size, (2) variation in fecundity amongindividuals, (3) overlapping generations, (4)geographic dispersion of populations, and (5)unequal numbers of males and females. Factors 2-5
are not plausible as cause of the observed lack ofcorrelation between the genetic diversity and actual
population sizes. Fluctuation in population size
represented by genetic drift severely reduces theeffective population size of population, which is the
harmonic mean of the actual number of individuals
in the last t generations [24]. For an endangered
species like E. nudicuspe nursing by many
conservation groups, the bottleneck effect is a
primary factor acting to reduce the effective size of
populations. The species rapidly establishes new
populations by conservation activities. Thus evenlarge population with N > 5000 are likely to have ahigh probability of being of recent origin and remain
influenced by the bottleneck effects.
Because E. nudicuspe only occurs in thewetlands along the river system, its distribution is
disjunct, and as a result, inter-river gene flow is
expected to be less than intra-river gene flow.
Sometimes wild animals occur the gene flow, butmany constructions, road, buildings and railways,
prevent long distant movements. E. nudicuspe ismainly pollinated by small sap chafers with limited
flight ability (personal obs.) and its seeds have no
specialized mechanism for long distance dispersal,
inter river gene flow is considered feasible due totheir geographical proximity.
But the hierarchical analysis of the population
gene structure of the species showed therelationships among the river system nor real
distance system. Especially B2 and B3 are placed
the east end of the distribution, near to E or D
population that are placed the center of the
distribution. D1, D2 and A2 populations are wellmanaged by three conservation groups. The
members of the groups sometimes visit anotherconservation area, so we guess that the people makethe gene flow among the different river system. On
the other hand, among strictly conserved area, C1,
C4, C5 F1 and F2, the gene flow aren’t almost
observed. Genetic distance is over 0.07 among the
populations. It was suggested that high Gst wasdependent on the isolated 4 populations. The
conservation activities cause unexpected gene flow.

C4

0.093

0.01
Genetic distance
0.011
C1

0.068

F1
0.057
0.042
F2

0.051
0.040
0.012

0.060

0.015
D2
0.005
D1
0.007
0.009
B2
0.007
0.0007

E2

C5
0.021

A2

0.001

0.0014E1
0.008 0.004
B3

0.001

Fig. 4. Phenogram using the neighbor joining
method based on Nei’s (1987) standard genetic
distance.

DISCUSSION
No significant correlation existed between any of
the genetic diversity parameters and the actual size
613

GEOMATE- Bangkok, Nov. 14-16, 2016[2] Tomita K, “The restoration Eriocaulon
nudicuspe distribution using a hearing method”,
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2004.
[3] Leakey R and Lewin R, Phylsiological Plant
Ecology, 3rd ed. New York: Springer, 1995.
[4] Ono T, “Geological, geographical andecological approach to the “Kinjo-hill”-changesof landscape since the Edo period-“, Kinjo
Gakuin Univ. Bulletin, Nature Sci. Vol. 9(2),2013, pp10-21.
[5] Naito Y et al., “Density-dependent selfing andits effects on seed performance in a tropical
canopy tree species, Shorea acuminate(Dipterocarpaceae)“, For. Ecol. Manage. Vol.
256, 2008, pp375-383.
[6] Naito Y et al., “Selfing and inbreeding
depression in seeds and seedlings of
Neobalanocarpus heimii (Dipterocarpaceae)”, J.
of Plant Res. Vol. 118, 2005, pp423-430.
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Inbreeding
depression
andoutbreeding
depression”, Evolution Vol. 45, 1991, pp622629.
[8] Templeton AR, “Coadaptation and outbreeding
depression”, Conservation Biology, Soule ME
Ed. Sinauer, Massachusetts, 1986, pp105-116.
[9] Price MV et al., “Pollen dispersal and optimal
outcrossing in Delphinium nelsoni“, Nature Vol.
277, 1979, pp294-297.
[10] Endangered Plant Survey Group, The Red Data
Book of Japanese Vascular Plants. Tokyo:
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[16] Nei M, Molecular evolutionary genetics. New
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[17] Saitou N & Nei M, “The neighbor joining
method: a new method for reconstructing

Generally each population is exposed to natural
selective pressure each area and adapts itself to it.
Therefore, if an artificial gene flow occurred, humandisappears an adaptive gene of to each habitat.
Sometimes annual fluctuation of environment selects
adaptive genes, another gene disappear the adaptive
gene by their reproductive ability when the selective
event did not occur. It is suggested that the artificial
gene flow decrease the gene diversity of endangered
species. It is suggested that we must not occur
artificial gene disturbance in order to conserve
endangered species. The conserved area should beprotected strictly. There were some populations withgene disturbance about E. nudicuspe observed.
Among these populations we cannot recover fromgene disturbance. We make a proposal that in order
to avoid the inbreeding depression high level geneflow management among the populations. Thatcause the uniformity of gene diversity, but there is
low inbreeding depression expected, so among
populations the individual has high fecundity.
CONCLUSION
We get three main conclutions.
1) There is no relationship between the real
distance and genetic distance.
2) There is a relationship between the conservation
activities and genetic structure.
3) There is no relationship between population sizeand genetic diversity.
Based on the conclusions, two proposals about theconservation activities
1) The conservation activities should be strictly
managed, because of unexpected genetic
disturbances.
2) We should make the networks among thehabitats where the genetic disturbance occurred,
the networks cause high level gene flow andavoid the inbreeding depression, because therecovering from genetic disturbance is very
difficult.
ACKNOWLEDGEMENTS
We wish to thank the members of Shimada
wetland conservation association for researchsupport and permission to work in the wetland. We
also thank the members of our laboratory for theirassistance in the field works.
REFERENCES[1] Ueda K, “Phytogeography of Tokai Hilly Land
Element I. Definition”, Acta Phytotax. Geobot.
Vol. 40, 1989, pp190-202.

614

GEOMATE- Bangkok, Nov. 14-16, 2016

phylogenetic trees”, Molecular Biology and
Evolution, Vol. 4, 1987, pp106-425.
[18] Maki M et al., “Genetic diversity andhierarchical population structure of a rare
autotetraploid
plant,
Aster
kantoensis(Asteraceae)”, American J. of Botany, Vol. 83,
Mar. 1996, pp296-303.
[19] Soltis DE & Soltis PS, “Genetic consequencesof autotetraploidy in Tolmiea (Saxifragaceae),
Evolution, Vol. 43, 1989, pp586-589.
[20] Ness BD et al., “Autotetraploidy in Heuchera
micrantha (Sazifragaceae)”, American J. of
Botany, Vol. 76, 1989, pp614-626.
[21] Wolf PG et al. “Chloroplast-DNA andelectrophoretic variation in diploid anautotetraploid
Heucheragrossularifolia”,

American J. of Botany, Vol. 77, 1990, pp232244.
[22] Shore JS, “ Tetrasomic inheritance and isozyme
variation in Turnera ulmifolia vars. elegance
Urb. and intermedia Urb. (Turnraceae)”,
Heredity, Vol. 66, 1991, pp305-312.
[23] Masuda M and Nishimura F, “Genetic diversityof restored endangered species, Penthorum
chinsense in the riverbed”, Int. J. of GEOMATE,
Vol. 10, 2016, pp1810-1814.
[24] Hartl, DL & Clark AG, Principles of population
genetics, 2d ed, Sunderland, MA, 1989, Sinauer.

615

GEOMATE- Bangkok, Nov. 14-16, 2016

Appendix Allel frequencies at 15 loci of 12 examined
populations of Eriocaulon nudicuspe.
population
locus

allele

A2

B2

B3

C1

C4

C5

D1

D2

E1

E2

F1

aco-1

N

19

22

28

30

28

28

14

19

30

30

30

30

a

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

b

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

aco-2

skdh

idh

mdh-1

mdh-2

mdh-3

acp-1

acp-2

acp-3

acp-4

F2

N

19

22

28

-

-

28

14

19

30

30

30

30

a

0.921

0.932

0.911

-

-

1.000

0.929

0.947

0.983

0.933

0.967

0.950

b

0.079

0.068

0.089

-

-

0.000

0.071

0.053

0.017

0.067

0.033

0.050

N

19

22

28

-

-

28

14

19

30

30

30

30

a

0.947

0.932

0.929

-

-

1.000

1.000

0.921

1.000

0.950

1.000

0.967

b

0.053

0.068

0.071

-

-

0.000

0.000

0.079

0.000

0.050

0.000

0.033

N

19

22

28

-

-

28

14

19

30

30

30

30

a

0.026

0.000

0.000

-

-

0.946

0.000

0.132

0.000

0.000

0.017

0.000

b

0.974

0.955

0.839

-

-

0.054

1.000

0.868

0.950

0.917

0.933

1.000

c

0.000

0.045

0.161

-

-

0.000

0.000

0.000

0.050

0.083

0.050

0.000

N

19

22

28

30

28

28

14

19

30

30

30

30

a

0.526

0.909

0.839

0.733

0.143

0.946

1.000

0.763

1.000

1.000

0.133

0.950

b

0.474

0.091

0.161

0.267

0.482

0.054

0.000

0.237

0.000

0.000

0.867

0.050

c

0.000

0.000

0.000

0.000

0.375

0.000

0.000

0.000

0.000

0.000

0.000

0.000

N

19

22

28

29

25

28

14

19

30

30

30

30

a

1.000

0.864

0.804

0.690

0.600

1.000

1.000

0.763

0.967

0.950

1.000

0.950

b

0.000

0.136

0.196

0.310

0.400

0.000

0.000

0.237

0.033

0.050

0.000

0.050

N

19

22

28

-

-

28

14

19

30

30

30

30

a

1.000

0.977

0.982

-

-

0.911

1.000

1.000

1.000

1.000

1.000

1.000

b

0.000

0.023

0..018

-

-

0.089

0.000

0.000

0.000

0.000

0.000

0.000

N

46

49

28

-

-

28

12

49

30

30

30

30

a

0.957

0.980

0.893

-

-

0.839

1.000

1.000

1.000

0.983

0.917

1.000

b

0.043

0.020

0.107

-

-

0.161

0.000

0.000

0.000

0.017

0.083

0.000

N

38

49

28

26

28

28

17

49

30

30

30

30

a

0.947

0.959

1.000

0.288

0.173

0.821

1.000

1.000

1.000

1.000

0.950

1.000

b

0.053

0.041

0.000

0.346

0.500

0.179

0.000

0.000

0.000

0.000

0.050

0.000

c

0.000

0.000

0.000

0.308

0.288

0.000

0.000

0.000

0.000

0.000

0.000

0.000

d

0.000

0.000

0.000

0.058

0.038

0.000

0.000

0.000

0.000

0.000

0.000

0.000

N

43

44

28

28

28

28

17

47

30

30

30

30

a

0.154

0.000

0.052

0.444

0.411

0.036

0.265

0.021

0.000

0.000

0.367

0.467

b

0.231

0.193

0.328

0.556

0.589

0.339

0.176

0.149

0.000

0.100

0.633

0.450

c

0.286

0.409

0.121

0.000

0.000

0.464

0.118

0.447

0.433

0.400

0.000

0.083

d

0.176

0.227

0.628

0.000

0.000

0.161

0.206

0.266

0.400

0.217

0.000

0.000

e

0.154

0.170

0.172

0.000

0.000

0.000

0.235

0.117

0.167

0.283

0.000

0.000

N

46

49

28

-

11

28

17

49

30

30

30

28

a

0.239

0.153

0.339

-

0.318

0.036

0.029

0.643

0.283

0.283

0.167

0.071

b

0.457

0.837

0.661

-

0.455

0.964

0.971

0.357

0.717

0.717

0.75

0.929

c

0.000

0.01

0.000

-

0.227

0.000

0.000

0.000

0.000

0.000

0.083

0.000

616

GEOMATE- Bangkok, Nov. 14-16, 2016

acp-5

pgi-1

pgi-2

pgi-3

pgm

mr

N

26

26

27

19

-

28

16

30

30

30

30

30

a

1.000

1.000

1.000

0.500

-

1.000

1.000

1.000

1.000

1.000

0.783

1.000

b

0.000

0.000

0.000

0.500

-

0.000

0.000

0.000

0.000

0.000

0.217

0.000

N

19

49

17

29

30

28

12

15

30

29

30

30

a

0.000

0.112

0.152

0.534

0.283

0.054

0.000

0.092

0.017

0.000

0.067

0.967

b

0.921

0.755

0.783

0.414

0.433

0.786

1.000

0.618

0.983

0.966

0.267

0.033

c

0.079

0.133

0.065

0.052

0.283

0.161

0.000

0.289

0.000

0.034

0.667

0.000

N

19

22

28

28

0

0

12

49

30

30

30

30

a

1.000

1.000

1.000

0.117

0.071

1.000

1.000

1.000

1.000

1.000

0.629

1.000

b

0.000

0.000

0.000

0.300

0.286

0.000

0.000

0.000

0.000

0.000

0.371

0.000

c

0.000

0.000

0.000

0.350

0.411

0.000

0.000

0.000

0.000

0.000

0.000

0.000

d

0.000

0.000

0.000

0.150

0.214

0.000

0.000

0.000

0.000

0.000

0.000

0.000

e

0.000

0.000

0.000

0.083

0.018

0.000

0.000

0.000

0.000

0.000

0.000

0.000

N

19

48

28

-

-

28

11

49

23

21

29

30

a

0.053

0.000

0.000

-

-

0.036

0.000

0.000

0.000

0.000

0.241

0.000

b

0.026

0.063

0.000

-

-

0.304

0.255

0.255

0.017

0.033

0.172

0.867

c

0.737

0.531

0.964

-

-

0.500

0.510

0.510

0.800

0.850

0.276

0.133

d

0.184

0.000

0.036

-

-

0.161

0.235

0.235

0.183

0.117

0.310

0.000

N

35

49

28

30

28

28

0,

34

30

30

30

30

a

0.348

0.061

0.000

0.150

0.536

0.679

0.000

0.000

0.000

0.000

0.000

0.000

b

0.485

0.776

0.893

0.850

0.464

0.286

0.542

0.468

1.000

1.000

0.017

1.000

c

0.167

0.143

0.107

0.000

0.000

0.036

0.458

0.456

0.000

0.000

0.850

0.000

d

0.061

0.020

0.000

0.000

0.000

0.000

0.000

0.076

0.000

0.000

0.133

0.000

N

19

22

28

30

30

28

12

49

30

30

30

30

a

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

1.000

b

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

0.000

N; No. of examination, -; non signal617

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THE ANALYSIS OF FLOODING CONDITION USING BY GIS DATA IN
THE THAILAND IN 2011
Masanobu Taniguchi1, Hiroyuki Ii1, Yuji Hara1 and Danai Thaitakoo2
1
Faculty of System Engineering, Wakayama University, Japan;
2
Faculty of Architecture, Chulalongkorn University, Thailand;

ABSTRACTThe purpose of this study is to clarify the cause of the flooding in Thailand that occurred in 2011. The Thailand
flooding in 2011 is analyzed by survey and GIS data. The flow condition in flooding was estimated by altitude
because water flows along to geographical features. The gradient is change at the lower site from Nakhon Sawan,
and the gradient at upper site from Nakhon Sawan is flat, and therefore it is estimated that the flooding isoccurred at Nakhon Sawan. The road is build higher than the other ground, so the road is shut the flooding water.
Therefore it is estimated that the flooded water could not flow out, there is the water between 3 months. The rain
runoff was analyzed by precipitation and GIS data. As a result, it was estimated that the rain runoff is bigger than
1.7 times of the river flow.
Keywords: Flood, Thailand, Precipitation, Runoff, GIS

INTRODUCTION
The flood was continued 3 months in 2011 in
Kingdom of Thailand. The total precipitation in theflood time is not particular big from 1985 to 2015.
As clear the cause of the floods, the river flow passof Chao Phraya River Basin is estimated. The waterpass is very complex by construction of agricultural
water. The flood extends to a fairly wide range. The
grasp of the flood area is very difficult, so it isdifficult to identify the study area. However it isneed to clear the cause of flooding. The study
purpose is clarify to cause of the flooding.

STUDY METHOD AND THE STUDY AREA
Fig.1 shows study area information. The data offlooding area was created using by locate offlooding point in Japanese Government Report[1][2].
The flood area data is roughly, it might contain anerror of several kilometers. The data is comparedwith altitude and the location data provide thedistance and slope. The average annual precipitation
is 1653mm/year, the annual precipitation in 2011 is
2240mm/year. The population of Thailand is 67
million people. The area of Chao Phraya River Basinis 160,400km2.
The usual flow rate in fine day is 718m3/sec. The
Chao Phraya River has 2 rivers, Nan, Ping, Chao
Phraya and Tha Chin. The Ping River and Nan River
meet in Nahon Sawan. The red triangle is the placeof flooding area in 2011. The aqua color line is canal,
the distribution of canal is between the Chao Phraya
River and the The Chin River.

Fig. 1

Study area information

The river, canal, and road data were created bytracing the river while looking at the map of Google
Earth. The local conditions was confirmed by GPS
photograph at fine day. Moreover the other placethat we not checked place was using by the Google
Street View[3]. The amount of rain runoff wascalculated by the monthly precipitation data and area
618

GEOMATE- Bangkok, Nov. 14-16, 2016

altitude (m)that was created by author’s GIS software using
space shuttle altitude data. The precipitation data is
opening to public by Japan Meteorological Agency.
The altitude data is published by Shuttle Radar
Topography Mission[6].

CHARACTERISTICS OF RAINFALL

500
475
450
425
400
375
350
325
300
275
250
225
200
175
150
125
100
75
50
25
0

0

Fig. 2 shows Time series of monthly
precipitation. The maximum of precipitation is
600mm, the minimum is 300mm. The precipitationin 2011 is 600mm, the value is not particularlyhigh[5].

Fig. 4

250

500

750
distance (km)1000

The altitude from Upper stream
Bangkok1250

toFig.4 shows the altitude from Upper stream to
Bangkok. The altitude of 600km distance from sea isabout 120m, the gradient of 600km distance fromsea is 0.0002. The gradient from 600km distancefrom sea to Nakhon Sawan is 0.0003. The height
difference from Nakhon Sawan to Bangkok is m per
250km. The gradient is calculated as 0.00012.

(mm/month)

700
600
500
400
300

NAHON SAWAN

200
100

Fig. 5 shows the map of around the Nakhon
Sawan. The area is the confluence of Ping Riverand Nan River. The purple color is Wetlands. The
red color line is old water pass that separated fromthe current river. The area is very flat, and all ofupper basin water is collected at this place. Thereis the raised bed river around the Nakhon Sawan asshown by altitude data. The flooding is occurred at
30m altitude, it is estimated that the water could
not flow to downstream.

BangkokFig. 2Phitsanulok

Nakhon Sawan

2015/7/1

2014/5/1

2014/12/1

2013/3/1

2013/10/1

2012/8/1

2012/1/1

2011/6/1

2010/4/1

2010/11/1

2009/9/1

2009/2/1

2008/7/1

2007/5/1

2007/12/1

2006/3/1

2006/10/1

2005/8/1

2005/1/10Chiang Rai

Time series of monthly precipitation(mm/month)

700
600
500
400
300
200
100

0
2011/6/1
Chiang Rai

Fig. 3

2011/7/1
Phitsanulok

2011/8/1
Nakhon Sawan

2011/9/1
BangkokMonthly precipitation of each observing
stationFig.3 shows monthly precipitation of eachobserving station[2]. The precipitation in June is alittle, the precipitation in August in Chang Rai is
about 2 times compared with the other place. It isestimated that the precipitation of Chang Rai is
important for the analysis of flooding, because
Chang Rai becomes water source of two big river[1].
However the observing station of Chang Rai has nodata some years. Moreover daily precipitation data
was not able to get.

Estimated flooding direction

Fig. 5619

The map of around the Nakhon Sawan

GEOMATE- Bangkok, Nov. 14-16, 2016

Estimated flooding
direction

Fig. 6

The picture at Nakhon Sawan

Fig. 6 shows the picture at Nakhon Sawan. The
red line shows water level of the flood in 2011. The
height is about 1.5m. The city is surrounded by therivers, so it is estimated that the water could not flowout.

altitude (m)

30

20

10

Fig. 800Fig. 7

10

20

30

40
50
distance (km)

60

70

80

The map of around the Chao Phraya Dam
Bhumibol Dam

The altitude from Nakhon Sawan to ChaiNat

Khuen Sirikit

122m
Ping River

Fig.7 shows the altitude from Nakhon Sawan to
Chai Nat. The height difference from Nakhon Sawan
to Chai Nat is 5m per 75km. The gradient iscalculated as 0.0006. The gradient changes big at
Nakhon Sawan before and after.

88m
Nan River
178m
146m
220m

47m
Canal

UPPER STREAM OF CHAO PHRAYA DAM

Chao Phraya Dam
81m
63m

Fig.8 shows the map of around the Chao Phraya
Dam. There is the dam across the Chao Phraya River.
The river breadth is about 245m. The dam can flows
max 3300m3/sec. The dam has 16 channel of 12.5m
breadth. And there is a diversion weir of agricultural
water upper stream of the Chao Phraya Dam. The
estimated flooding direction is from river to ground.
It is estimated that the water flows over the bank of
the river at the flooding point, because the river
altitude is higher than the lower land point.
Fig.9 shows the overview of breadth of the Chao
Phraya River. As the gradient is a very little inwhole, the river breadth is important for the river
flow. The gradient is same or a little change between
some hundred kilometers. It is estimate that the

167m
dam

Fig. 9113m

Bangkok

The overview of the Chao Phraya River
breadthbreadth is almost decide the flow rate. The somebreadth is narrow at the main stream in the Chao
Phraya River. Moreover it is estimated that the
Chao Phraya Dam has stopped the flow of water.
And it is thought that the water level at the upper
dam became higher at least a little.

620

GEOMATE- Bangkok, Nov. 14-16, 2016

INFULUENCE OF DAMS
Table 1 shows the volume of dam. The total ofvolume is 24 billion m3. Bhumibol Dam is biggest in
the Chao Phraya River Basin. Khuen Sirikit is 9.5
illion m3.

Phraya Baisn. Phitsanulok is located in middle
stream of the Chao Phraya River, the Chao Phraya
River meet at Nakhon Sawan, all of the basin’s
water corrected in the Nakhon Sawan. It is need to
calculated the amount of rain runoff.

Table.1 The volume of dam

70,000,000,000
60,000,000,000
50,000,000,000
40,000,000,000
30,000,000,000
20,000,000,000
10,000,000,000
03

Dam Name
Bhumibol Dam
Khuen Sirikit
Other
Total

Volume(m )13,500,000,000
9,500,000,000
1,700,000,000
24,700,000,000

Rain runoff(m3)Chiang Rai Chiang Mai Phitsanulok
2010

ANALISYS OF RAIN RUNOFF

Nakhon
Sawan

NakhonBangkok
Ratchasima

2011

Fig. 10 The amount of rain runoff of four months
Fig. 10 shows the precipitation of between fourmonths. The area is calculated by Thiessen Law,
and to calculate the amount of rain runoff is need thearea. The amount of rain runoff is calculated bymultiplied by the amount of rainfall in the area. The
area is divided by the river basin, and the data of theriver basin is created by the altitude data. The flood
is occurred in July, and the flood is continued
between 3months, the amount of the precipitation issummed from June to September. The precipitationof the period tends to bigger in a year. The

2011

Fig. 10 shows the amount of rain runoff of fourmonths. The amount of rain runoff was comparedwith in 2010 and in 2011. The monthly precipitation
changes very little by a month. The anount of rain
runoff in 2011 ara increase over 10 billion m3
compared with 2010 in Chang Rai and in
Phitsanulok. In particular, In Phitsanulok, theamount of rain runoff increase 25billion m3
compared with in 2010. In fact, the lowerstream ofPhitsanulok at Nakhon Sawan was flooding, the city
was big damaged.
5,000
4,500
4,000
3,500
3,000
2,500
2,000
1,500
1,000
500
0

Chiang Rai
1509mm17171 km2

Chiang Mai
921mm
41188 km2

Phitsanulok
1261mm49143 km2

Rain runoff(m3/sec)

Chiang Rai Chiang Mai Phitsanulok Nakhon
NakhonBangkok
Sawan Ratchasima

2010

Nakhon Sawan
797mm
40959 km2

2011

Fig. 11 The amount of rain runoff per second
Nakhon Ratchasima
721mm
5777 km2

Fig. 11 shows the amount of rain runoff per
second. The rain runoff was compared with theriver flow rate. The actual flow rate is unknown. To
collect the daily precipitation data is difficult. The
daily rain runoff data was calculated by the monthly
precipitation data between 4 months and the basin
area. The Chao Phraya Dam is possible to flow
3300m3/sec. The amount of rain runoff at upper the
Chao Phraya Dam is 5800 m3/sec, the rain runoff is
1.7 times from the Chao Phraya Dam flow.

Bangkok
1251mm
38073 km2

Fig. 10 The precipitation of four months
precipitation in Phitsanulok and in Chang Rai is overthan 1000 mm, and moreover the area is large. So
the amount of rain runoff is biggest in the Chao621

GEOMATE- Bangkok, Nov. 14-16, 2016

level at the upper Chao Phraya dam became higher
at least a little. The road is higher than the otherground. Therefore it is estimated that the water isdifficult to flow over the road. It is estimated that theflooded water could not flow out, there is the water
between 3 months.

IMPACT OF THE ROAD
Fig. 12 shows the overview of the road. The
flooding area is devided at National load No.1 andNo.346 located near Bangkok shown in Fig.1. The
road is higher than the other ground. Therefore it isestimated that the water is difficult to flow over theroad. It is estimated that the flooded water could not
flow out, there is the water between 3months.

REFERENCES[1] Land, Infrastructure and Transportation
Ministry,
http://www.mlit.go.jp/river/kokusai/disaster/thailand/thailand_flood_111111.pdf.
[2] Japan
Meteorological
Agency,http://www.data. jma.go.jp/gmd/cpd/monitor/[3] Google Map, http://www.google.co.jp[4] Masanobu Taniguchi, and Hiroyuki Ii ,”Distribution characteristics of the Annual
nitrogen load in Yamato River Basin in 2011”,
International Journal of GEOMATE, June, 2016,
Vol. 10, Issue 22, pp. 2043-2049 , Japan

Fig. 12 The overview of the road
CONCLUSION
The study purpose is clarify to cause of theflooding. The Thailand flooding in 2011 is analyzed
by survey and GIS data. The data of information ofwater control was collected as location data. The
gradient changes big at Nakhon Sawan before andafter, at the lower Nakhon Sawan, and the gradient is
a very little, and therefore it is estimated that theflooding is occurred. And it is thought that the water[5] Kazuyuki Mizumura, Hydrosphere hydrology,
1998.
[6] Shuttle Radar Topography Mission, http://
e4ftl01.cr.usgs.gov/SRTM/SRTMGL3.003/200
0.02.11/622

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

UTILIZATION OF TOFU LIQUID WASTE GENERATED FROMANAEROBIC PROCESSING IN COMPOST PREPARATION
M. Faisal
Department of Chemical Engineering, Syiah Kuala University, Banda Aceh, Indonesia

ABSTRACTThis study aims to identify the ability of tofu liquid waste generated from anaerobic processing to act as a
microorganism source to help expedite compost decomposition. In this research, composting was performed on 2
kg of rice husks to which tofu liquid waste was added in various volumes (500–2000 mL) and 1 kg of cow
manure. This research used a nonfactorial completely randomized design with five treatments and four
repetitions, which made a total of 20 unit trials. Changes in pH, humidity (rH), C composition, N, C/N ratio, andnumber of colonies were analyzed to observe the compost quality. The data collected were then analyzed using
SPSS 16. The research findings showed that increasing the volume of tofu liquid waste has a significant impact
on composting temperature, and all compost outputs were within proper compost criteria. Increasing tofu waste
volume can improve the population of microorganisms, which accelerated the decomposition process of organic
matter. The highest number of bacterial colonies, 73 colony forming units, was found on 1500-mL treatment. At
this condition the values obtained were 17.94%, 0.9%, and 20.1 for C, N, and C/N ratio, respectively.
Keywords: Compost, Tofu Waste, C/N Ratio, Bacteria Colonies

quality that meets effluent regulations. One of themethods to utilize the waste is making it into a
microorganisms’ source in the composting process.
Composting technology adoption is one of thealternatives that can reduce the amount of waste
production [5]. Composting is one of the methods
that have proven to be efficient and effective inwaste processing, from both economic andenvironmental points of view [6]. Many studies oncomposting using various types of microorganisms
have been done with relatively effective results[7],[8]. The success of compost production isdetermined by the ability of decomposer
microorganisms to decompose organic matter.
Mancebo and Hettiaratchi [9] conducted an intensive
study on assessment of the composting dynamics ofdifferent organic residues (i.e., food wastes,
dewatered sludge, sawdust, and herbal). The aerobic
composting process is a cost-effective technique forthe treatment and sanitization of biodegradable solid
wastes fractions [10-13]. Composting can also
effectively improve soil condition and farm productquality [14]. Nowadays, different types of
microorganisms are added to accelerate thecomposting process, one of which is EM4, whichcontains a mixed culture of microorganisms.
The objective of this research was to identify theability of tofu liquid waste produced from the TSA
process to act as a microorganisms’ source to help
accelerate the decomposition process of composting.
It was expected that the mixed-cultures bacteria
contained in the tofu waste could accelerate thedecomposition process of the compost.

INTRODUCTION
Tofu is a traditional oriental food and is typicallyprocessed by grinding and boiling of soybeans. The
tofu industry is one of the industries that use a large
amount of water in its processes. The water used in
the process produces a large amount of tofu waste.
Without a good handling process, tofu waste
containing a high proportion of organic compounds
may cause various adverse impacts, such as waterpollution, various diseases, bad smell, increased
mosquito population, and degrading the aesthetics ofthe environment. Liquid waste disposed to a water
body without prior processing may also kill thebiotic community in the water, including
microorganisms, which have important roles inmanaging biological balance in the water. Small tofu
processing industries usually dispose of their waste
into the sewage system and do not go throughappropriate processing [1],[2]. Some researchers
have utilized liquid waste from the tofu industry assources of methane and hydrogen gas [3],[4] bybiological degradation (anaerobic process). Aftergoing through an anaerobic process, tofu liquid
waste still contains mixed cultures of bacteria andorganic matter. In previous research, Faisal et al. [4]used a thermophilic stirred anaerobic (TSA) reactor
to process tofu liquid waste. The chemical oxygen
demand value of the waste generated by the TSA
process was in the order of 1900–2000 mg/L. Thishigh organic matter could not be discharged to theenvironment, Thus, further processing was required(for example, aerobic, membrane) for proper waste
623

GEOMATE- Bangkok, Nov. 14-16, 2016

analysis of variance using SPSS 16 showed that the
tofu waste volume treatment had a significant impact(P = 0.001 < 0.05) on the compost. The observation
results revealed that the average composting
temperature during each treatment showed irregulartemperature fluctuation and this was almost the samein each treatment.

RESEARCH METHOD
The materials used to make compost in thisresearch consisted of rice husks, manure (thematured type), and tofu liquid waste produced from
TSA reactor processing. The characteristic of tofu
waste has been described previously [2].
Trial Design

ControlTemperature (oC)This research used a nonfactorial completely
randomized design with five treatments and four
repetitions, which made a total of 20 trial units.
Types of treatment: P0: 2 kg rice husks + 1 kg
manure; P1: 2 kg rice husks + 1 kg manure + 500
mL tofu liquid waste; P2: 2 kg rice husks + 1 kg
manure + 1000 mL tofu liquid waste; P3: 2 kg ricehusks + 1 kg manure + 1500 mL tofu liquid waste;
P4: 2 kg rice husks + 1 kg manure + 2000 mL tofu
liquid waste.
Data analysis was conducted using SPSS 16,
where if the analysis of variance showed asignificant difference, it would be followed by the
Smallest Real Difference Test at the 5% level.

39

500 mL

37

1000 mL

35

1500 mL

33

2000 mL

31
29
27
25
1 5 9 13 17 21 25 29 33 37 41 45 49 53 57
Time (Day)

Fig. 1 Temperature as a function of time.
Figure 1 shows that the highest composting
temperature of 34.75 °C occurred with the 1000-mL
treatment during the first week of the composting
period. In contrast, the lowest temperature of
26.75 °C occurred on the 1500-mL treatment during
week 21. The observation results also revealed that
adding tofu waste in various volumes caused
significant temperature increases during thecomposting period.
The maximum temperature of the compostingprocess involving thermophilic microorganisms is
45–59 °C, where at that point the compostingprocess will be better [15]. In this study, the initialhighest composting temperature was only 34.75 °C,
which indicated that all treatment variations did notreach the required temperature for thermophilic
microorganisms to grow and develop. In other words,
mesophilic microorganisms played a role in thiscomposting process. This was presumably caused by
the high humidity of the pile, which prevented
maximum heat production. This caused the compost
to be unable to reach the optimum temperature
required for thermophilic microorganisms to grow,resulting in a slow composting process. The compost
temperature declined after the first week as
microorganisms’ activity to decompose organic
matter reduced. The temperature decline could alsomean that the compost had matured andmineralization process of nutrients had taken place.

Experimental
Preparation of compost
The materials used to make the compost wererice husks and manure, at a ratio of 2:1. One (1) kgof trash-free dry rice husks was mixed with 2 kg ofmanure. The manure used in the process was
matured manure that had characteristics of low
temperature, less strong smell, and dry. All bacteria(tofu waste) were fed into the composter (with acapacity of 30 L and equipped with a temperature
indicator). The temperature, pH, and relative
humidity (rH) of the composting were observed on adaily basis for 60 days. The observation was limited
to the C/N ratio, C organic, total N, and the numberof decomposer microorganism colonies. The
analyses followed the procedure given in Standard
Methods for the Examination of Water andWastewater. Determination of total nitrogen was
done based on the total Kjeldahl nitrogen content
and carbon was measured by gravimetric analysis.
RESULTS AND DISCUSSION
Temperature Changes
Temperature is one of the key indicators in compost
making. Heat is generated by microorganisms’
activity when decomposing organic matter.
Temperature can also be used to find out how well
the composting process is and to what extent
decomposition has taken place. The results ofCompost Characteristics (C/N Ratio, Organic C,
and Total N)
On day 60, chemical analysis to find out compost624

GEOMATE- Bangkok, Nov. 14-16, 2016

quality was carried out. The compost testing
consisted of C, N, and C/N ratio analyses. The
results of laboratory analysis of the matured
compost’s chemical contents are shown in Table 1.
Tables 2 and 3 show the standard compost criteria
according to the Indonesian National Standard (SNI)and Regulation of the Minister of Agriculture No.
2/Pert/HK.060/2/2006, respectively [16].

treatment requires longer decomposition to reach thelevel where the C/N ratio is <20.
As shown in Table 3, however, all treatments tothe tofu waste compost with various volumes met
the criteria of good solid compost. The analysis
results also showed that organic C and total N
contents on all treatments met the criteria of goodcompost for agriculture.
The high C/N ratio found in the 2000 mL volume
treatment was presumably due to too much tofu
waste being added to the compost, causing low
temperature and a slow decomposition process. In
contrast, the low C/N ratio in the control treatment
was due to temperature rise, causing rapid
microorganism effectiveness. The higher thetemperature, the lower the C/N value will be [20].
The decline in C/N ratio was due to the increased
microorganisms’ activity during the organic matter
break down process. The pile containing too little
nitrogen would not produce heat for a rapid break
down of matter, causing the C/N value to stay high.

Table 1 Compost’s chemical contents.
C
Treatment organic
(%)

N
total
(%)

Ratio
C/N

TemperH
rature
(%)
(°C)

pH

Control

20.52

1.09

18.73

29.9

0

7

500 mL

19.03

0.94

20.42

28.7

12.4

6.7

1000 mL

18.48

0.87

21.09

29.9

20.2

6.5

1500 mL

17.94

0.9

20.1

29.1

44

5.6

2000 mL

17.51

0.7

25.08

29.9

53. 6

5.3

Acidity (pH) and Humidity (rH)
1.Acidity (pH)The results of data analysis using SPSS showedthat the influence of tofu waste volume on compost
pH was significant, where P < 0.05, thus H0 was
rejected, which means that the average acidity (pH)of each volume treatment applied to the tofu waste
was different. The pH change in the compostingprocess varies vastly [21] depending on the types ofcomposting materials and the microorganisms in it.

Table 2 Compost characteristics according to theIndonesian National Standard (SNI-19-7030-2004).
C
N
Limit organic total
(%)
(%)
Min.
9.8
0.4
Max.
32
–

Ratio
C/N
10
20

Temperature
(°C)
–
30

rH
(%)

pH

–
50

6.8
7.49

Table 3 Compost characteristics according to theRegulation of the Minister of Agriculture No.
2/Pert/HK.060/2/2006.
N
total
(%)

Ratio
C/N

≥12

0.4

10 ± 25

Solid

TemperH
rature
(%)
(°C)
–

8
pH
6

13 ± 20 4 ± 8

pHCContents organic(%)

The C/N ratio is a comparison between carbon
and nitrogen. The C/N ratio is widely used as anindicator of compost maturity [17]. The C/N ratio isused to identify the presence of microorganisms inan area because nitrogen is absorbed by plants anddead microorganisms leave carbon sediment.
A low C/N ratio of between 4 and 10:1 generally
comes from the sea. Plants with wood vessels onland will provide sediment (organic matter pile) with
C/N ratio <20 [18]. According to Setyorini et al.[19],the C/N ratio of good compost is 10–20, dependingon the raw materials and humidity level.
Based on Table 1, the C/N characteristics ofcompost according to the SNI on the control
treatment and the 1500-mL treatment with C/N ratio
value were 18.73 and 20.1, respectively. Table 1
also shows that the C/N ratio from the highest
compost treatments, i.e., in volume 2000 mL, was anaverage of 25.08 and the content of the N element
was 0.7%, which was also the lowest of alltreatments. We assume that the 2000 mL compost

4
2
0
500 mL 1000 mL 1500 mL 2000 mL
Treatment

Fig. 2 Influence of waste volume on pH of compost.
Figure 2 shows the influence of waste volume on
pH of compost. Compost acidity criteria are: very
acidic <4.5, acidic 4.5–5.5, semi-acidic 5.6–6.5,
neutral 6.6–7.5, semi-alkali 7.6–8.5, alkali >8.5. In
Figure 2, the 1500 mL and 1000 mL volumes were
included in semi-acidic, the 500 mL and control
were included in the neutral criteria. Matured
compost usually has a pH very close to neutral. The
pH of optimum composting ranged between 5.5 and8.0. The pH of the 2000-mL treatment was includedin acidic criteria as the activity of the
microorganisms was to decompose organic matterand to produce simple organic acids. This signified
that the composting process on that particulartreatment was still continuing.
625

GEOMATE- Bangkok, Nov. 14-16, 2016

The composting process of organic matter canoccur in a large pH range. The composting alone
would cause changes to the organic matter and thepH. Although the pH of several treatments involving
tofu waste was still in the acidic category, according
to the standards in Table 3, all of the treatments with
tofu waste met the criteria of compost used inagriculture, as the pH of the compost was higher
than 4 and close to neutral.

Number of Colony (CFU)

related to the value of C/N for that particulartreatment. In the 1500-mL treatment, the bacteria
had sufficient energy sources.

rH (%)

2.Humidity (rH)
Humidity plays an important role in themetabolism process of microbes and indirectly
influences oxygen supply. Microorganisms canutilize organic matter if the organic matter isdissolved in water. Sufficient humidity is required
during the composting process [22]. High levels of
humidity are harmful for efficient compostingdevelopment because of decreasing the air void
volume available for oxygen movement [22].
The results of data analysis showed that the
influence of tofu waste volume on compost rH wassignificant, where the value of P < 0.05, so that H0
was rejected. In other words, the average humidity(rH) of each tofu waste volume treatment was
indeed different.

40
30
20
10
0
500 mL

1000 mL 1500 mL 2000 mL
Treatment

Fig. 4 the influence of various tofu waste
volumes on the number of bacteria colonies.
Although there was a difference between theanalysis of variance and the number of bacteriacolonies in each treatment, adding tofu waste in thecomposting process gave significant results. This isconsistent with Hindersah’s study [20], which foundthat adding tofu waste and bioactivator (cowmanure) can increase the quality of compost
microbiology. It is explained further that tofu waste
can increase the total population of bacteria, fungi,
Azotobacter sp. and phosphate-solubilizing bacteria.
The basic components required as nutrients for
microorganisms are carbon, nitrogen, and water.
Microorganisms will use C compounds as an energy
source to breed and use N to synthesize protein.
During the composting process, many fungi andbacteria are responsible for the degradation ofcomplex polymeric substrates, such as pectin, lignin,cellulose, and hemicellulose [8]. Li et al. [23]reported that, in general, the most efficientlignocellulose degraders are fungi due to the factthat their mycelial structure has a competitive
advantage over bacteria.
The species of bacteria found in the compost on
each treatment remain unknown. It was assumed thatthe decomposer bacteria that played a role indecomposing the organic material in the tofu waste
compost were the lactate acidic bacteria(Lactobacillus sp.). Bacteria of this type arecommonly found in composting material. Bacteria
Lactobacillus sp. were able to reduce dangerous
microorganisms and to decompose organic matter
rapidly.

60
50
40
30
20
10
0
500 mL

50

1000 mL 1500 mL 2000 mL
Treatment

Fig. 3 Influence of waste volume on rH of compost.
Figure 3 shows the influence of tofu waste oncompost rH. It can be seen that the smaller thevolume of tofu waste in the composting process, thelower the humidity. Figure 3 also shows that the
500-mL treatment had humidity below the standard
matured compost. The 1000-mL, 1500-mL, and2000-mL treatments met the criteria of propercompost, as indicated in Table 3.
Number of Bacteria Colonies
Figure 4 shows the influence of various tofu
waste volumes on the number of bacteria colonies in
the compost. Analysis of variance showed thattreatment of adding tofu waste did not have asignificant influence (P = 0.744 > 0.05) on thenumber of microorganisms, especially bacteria. In
other words, the number of bacteria colonies in eachtreatment was the same or no different from one
another.
The maximum number of bacteria colonies was
mostly found in the 1500-mL treatment. This was

CONCLUSIONS
Microorganisms found in tofu waste generatedusing a TSA reactor can be used as decomposers in
the making of rice husk compost. Increasing tofu
waste volume can increase the population of626

GEOMATE- Bangkok, Nov. 14-16, 2016

microorganisms, which accelerates the organic
material degradation process. All treatments of tofu
waste compost involving various volumes meet thecriteria of good solid compost. The analysis results
also showed that organic C and total N contents inall treatments meet the criteria of good compost for
agriculture. The highest number of bacteria colonies
was mostly found in the 1500-mL treatment. Further
research with various compost quality analyses (e.g.,
P, K, trace elements) will support this preliminary
study.

[7] Chen Y, Zhou W, Li Y, Zhang J, Zeng G,
Huang A, Huang J, “Nitrite reductase genes asfunctional markers to investigate diversity of
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May 2014, pp. 4233-4243.
[8] Ventorino V, Parillo R, Testa A, Viscardi S,
Espresso F, Pepe O,”Chestnut green waste
composting for sustainable forest management:
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Jan 2016, pp.168-177.
[9] Mancebo U, Hettiaratchi JPA, “Rapid
assessment of methanotrophic capacity ofcompost-based materials considering the effectsof air-filled porosity, water content anddissolved organic carbon”, Bioresour. Technol.
Vol.177, Feb. 2015, pp.125-133.
[10] Komilis D, Evangelou A, Voudrias E,“Monitoring and optimizing the cocompostingof dewatered sludge: a mixture experimental
design approach”, J. Environ. Manage. 92 (9),Sep. 2011, pp. 2241-2249.
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and lime addition enhances the rate andefficiency of decomposition of food waste
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100, Jul. 2009, pp. 3321-3324.
[13] Zhou Y, Selvam A, Wong JW, “Evaluation of
humic substances during co composting of foodwaste, sawdust and Chinese medicinal herbal
residues”. Bioresour. Technol., Vol.30, No.168,
Sep. 2014, pp. 229-234.
[14] Mandal A, Patra AK, Singh D, Swarup A,
Masto RE, “Effect of long-term application ofmanure and fertilizer on biological andbiochemical activities in soil during cropdevelopment stages”, Bioresour. Technol. Vol.
98 , Dec. 2007, pp. 3585–3592.
[15] Tom LR, “Municipal solid waste composting:biological processing”, Center for theEnvironment, Cornell Waste Management
Institute, 1993, pp. 1-6[16] National Standardization Agency, Compost
Spesification from domestic organic wastes,
Indonesian National Standard 19-7030-2004.

ACKNOWLEDGEMENTS
This work was supported by the Ministry of
Research, Technology, and Higher Education ofIndonesia with a research grant for International
Research Collaboration and Scientific Publication.
The authors also thank Ms. Afriyani, who helped in
the data analysis of this study.

REFERENCES[1] Faisal M, Machdar I, Mulana F, Daimon H,“Potential renewable energy from tofu
processing waste in Banda aceh city, Indonesia.
Asian J. Chem.”, Vol. 26, No.18, Oct. 2015, pp.
6601-6604.
[2] Faisal M, Mulana F, Gani A, Daimon H,“Physical and chemical properties ofwastewater discharged from tofu industries in
Banda Aceh city, Indonesia”, Res. J Pharm. Bio.
Chem. Sci., Vol. 6, No. 4, Jul. 2015, pp. 10531058.
[3] Lay CH, Sen B, Huang SC, Chen CC, Lin CY,“Sustainable bioenergy production from tofuprocessing wastewater by anaerobic hydrogen
fermentation for onsite energy recovery”,
Renew. Energ., Vol. 58, Oct. 2013, pp 60-67.
[4] Faisal M, Gani A, Mulana F, Daimon H,“Treatment and utilization of industrial tofu
waste in Indonesia”, Asian J. Chem, Vol. 28,
No.3, Feb. 2016, pp. 501-507.
[5] Mohee R, Boojhawon A, Sewhoo B,
Rungasamy S, Somaroo GD, Mudhoo A,“Assessing the potential of coal ash and bagasseash as inorganic amendments duringcomposting of municipal solid wastes”, J. Envi.
Mana., Vol. 159, Aug. 2015, pp. 209-217.
[6] López R, Giráldez I, Palma A, Díaz MJ,“Assessment of compost maturity by using anelectronic nose”, Waste Mana., Vol. 48, Feb.
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GEOMATE- Bangkok, Nov. 14-16, 2016[17] Bernal MP, Paredes C, Sanchew-Monedero MA,
Cegarra J, “Maturity and stability parameters ofcomposts prepared with a wide range of organic
wastes”, Bioresour. Technol., Vol. 63, Jan. 1998,
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[18] Raj D, Antil RS, “Evaluation of maturity andstability parameters of composts prepared fromagro-industrial wastes”, Bioresour. Technol.,Vol 102, No.3, Feb. 2011, pp.2868-2873.
[19] Setyorini DR, Saraswati, Anwar EK,“Compost”, Research Center for Agricultural
Technology, DI Yogyakarta, 2003.
[20] Hindersah R, Hernanto Y, Joy B, Mulyani, O,“Utilization of tofu waste in composting ofhousehold waste for increasing quality ofcompost microbiology", Agrinimal, 2011, pp.
15-21.

[21] Juárez MFD, Prähauser B, Walter A, Insam,
H ,Franke-Whittle IH, “Co-composting of
biowaste and wood ash, influence on a
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[22] Ahn H, Richard T, Glanville T, “Optimum
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[23] Li H, Xu X, Chen H, Zhang Y, Xu J, Wang J,
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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

COMPARATIVE ANALYSIS OF NEW GROUND MATERIALAND EMBANKMENT CONSTRUCTION METHODS INCONSIDERATION OF RECYCLING BY LCA
Hideyuki Ito1, Koichi Yamanaka 2, Hideo Noguchi 3, Takahiro Fujii 4 and Kunio Minegishi5
1,2,4,5
College of Science and Technology, Nihon University, Japan;
3
Tsuchiura City Hall, Japan
ABSTRACTThe cutting and embankment construction method, which is a conventional embankment construction
method, would possibly cause grand subsidence or landside at a location of a soft ground or with the risk oflandside. From these points of view, the construction method employing a new ground material such as EPS(Expanded Polystyrene) is effective for construction at such a location with soft ground or the landslide risk
because of its light weight and workability. However, such new materials have not much construction examples,
expect a few existing researches such as Ito et al. (2014) conducing comparative analysis in terms ofenvironmental load and cost in view of life cycle assessment (LCA). Thus, the accumulation of researches on the
new materials is still not enough. The purpose of this research is to extract what to be improved and solved in
new ground materials by analyzing various environmental loads and life cycle costs, which arise from the life
cycle in case of various embankment construction methods in consideration of whether they have used new
ground materials and recycle of waste materials. As a result, the research demonstrated a possibility of reducing
the environmental loads or life cycle costs by employing recycling in the various embankment constructions
methods.
Keywords: Life Cycle Assessment, Lightweight Geomaterial Mixed with Expanded Polystyrol Beads, Expanded
Polystyrol Construction Method, Foamed Waste Glass Construction Method, Life Cycle Cost

There have been a number of conventional
studies examining the new geomaterials, but few
have provided examples of their use in construction
work or performed LCA analyses. Conventional
comparative analyses of environmental load andcosts from the LCA perspective include those
performed by Ito et al. [1] and Ochiai et al. [2], who
introduced the possibility of using waste in mixed
geomaterials, but there has been an insufficientnumber of studies on the use of new geomaterials
containing recycled embankment materials.
These present researches aim to identify areas for
improvement and the problems associated with theuse of new geomaterials. Thus, in this research, weestimated CO2 and air pollutant emissions generated
by SOx and NOx, by comparatively analyzing the
LCCs of four construction methods throughout thelife cycle of the materials, from raw materials
collection to construction, and from usage todisposal. For our analyses, we assumed an earth
filling design for a mountainous area with thepossibility of land slide. We analyzed four methods
in the construction of embankments in mountainous
areas, including the conventional cut-and-fill method,the expanded polystyrol construction method using
the new EPS geomaterial, the lightweight
embankment construction method with EPS beads,
and the foamed waste glass construction method
using embankment material with recycled waste.

INTRODUCTION
Among the various embankment construction
methods, the conventional cut and fill method cancause ground subsidence and landslides on soft
ground and in landslide-prone areas. To deal withthis issue, a construction method has been developed,
using a new lightweight and workable composite
geomaterial mixed with expanded polystyrene (EPS),
which is believed to be effective for construction on
soft ground and landslide-prone areas.
Using Life Cycle Assessment (LCA) and Life
Cycle Cost (LCC), Ito et al. [1] have analyzed theenvironmental load and costs of embankment
construction using the cut-and-fill and newly
developed EPS construction methods. The results
show potential for reducing the environmental loadby recycling embankment materials. Ochiai et al. [2]have summarized the added value and physicalproperties of various mixed geomaterials and haveclassified the constituent materials. They also used
LCA to analyze embankment materials composed ofrecycled tires, with respect to the manufacturing
process for the materials alone. Inazumi et al. [3]also assessed the recycling of construction sludge
generated from embankment works. Onizuka et al.
[4] have described the engineering characteristics offoamed waste glass material and provided usefulexamples of its applications.
629

GEOMATE- Bangkok, Nov. 14-16, 2016of specific applications. Hence, foamed waste glass
is used in a wide range of applications including
civil engineering, greening of slopes and rooftops,
agriculture, water purification, and heat insulation.
This material is lightweight, water permeable, water
retentive, fire resistant, and a good thermal insulator.
Also, we set a functional unit that provides alogical basis for comparing the environmental
performance of alternatives for applying LCA tothese 4 construction methods. We defined the target
road condition (2 lanes, 7m wide and 1m long) as afunctional unit as shown in the Fig.2. In addition, we
hypothesized that the inclined angle of the mountainand road slope is 35 degree respectively (Table 1).

METHOD AND SYSTEM DETAILSIn this research, Mineoka area located in thesouthern part of Chiba prefecture was selected as acase research area as well as previous research [1]because landside control works are conducted in
mountainous roads (Fig.1). For comparative LCA
analysis, we assumed conventional method,
lightweight geomaterial mixed with expanded
polystyrol beads and expanded polystyrol
construction method and the foamed waste glass
construction method using embankment material
with recycled waste for constructing mountainous
roads.
Based on an embankment design for amountainous area with the possibility of land slide,
we focused on the weight savings and recycled
embankment materials in our analyses of theconventional cut-and-fill method, the EPS methodwith new geomaterials, the lightweight embankment
construction method with EPS beads, and the
foamed waste glass construction method using
recycled waste glass as embankment material.
In the EPS method, EPS blocks are stacked asembankment materials and are integrated bydedicated clamps. When stacked, these ultralightweight embankments are characterized by theircompressive resistance, durability, and independent
stack design.
In the lightweight embankment construction
method with EPS beads, lighter earth is used,comprising EPS mixed with earth and sand. Thismethod is effective for use in earth fills on soft
ground and in landslide-prone areas due to its
capability of reducing the applied load on the groundmore effectively than ordinary earth and sand.
Foamed waste glass is a porous embankment
material manufactured by pulverizing, burning, andfoaming recycled waste glass. The specific gravity
and degree of water absorption can be controlledduring manufacturing according to the requirements

Fig. 1

Location of case research area.

Fig. 2Functional unit.

Table 1 Set condition of each construction method.

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GEOMATE- Bangkok, Nov. 14-16, 2016

SYSTEM BOUNDARYMETHODSANDRECYCLING

The system boundary of the foamed waste glass
construction method using embankment material
with recycled waste is shown in Fig.3. Foamed
waste glasses are produced by recycled glass in theplant, and then they are leveled and compacted. At
the waste phase, used embankment materials will berecycled. We assumed that the method used recycled
embankment material of foamed waste glassmaterials collected after the roads had been
demolished. We established distances based on thelocation of factories, etc. located around the
Mineoka mountain district in Chiba Prefecture,
where embankment work is often conducted.
To calculate the total amount of air pollutants
and CO2 emitted by each method, we set the CO2,
SOx and NOx unit (Table 3) and cost unit for eachmaterial (Table 4). These units were developed
based on data from sources such as the LCA
guidelines for building [5], IDEA (Inventory
Database for Environmental analysis) [6], the LCA
database developed by the Life Cycle Assessment
Society of Japan [7], the database of the Express
Highway Research Foundation of Japan [8] and the

The system boundaries of the conventional cutand-fillmethod, theexpandedpolystyrol
construction method using the new EPS geomaterial,
the lightweight embankment construction methodwith EPS beads were set up same condition ofprevious research [1].
To consider the recycling impacts, we performed
four embankment tasks and assumed that the roadwould be reconstructed every 100 years after its
initial construction for 400 years. For the EPS andwaste glass cases, as used in system boundaries oflightweight soil mixed with EPS beads, we
calculated the CO2, air pollution, and LCCs for all
life cycles with and without recycled materials. As
recycling methods for most embankment materials
are still in the research and development phases, webased our analyses on a method used in the industry
for recycling embankment materials, as identified byresults from a questionnaire and interviews with 10
business operators. Table 2 shows the recycle
conditions of each construction method.

Table 2 Recycle conditions.
Construction method

Recycle method(first time)

Recycle method(after second time)
Banking material used for cut-and-fill are recycled
by mixing cement

Cut-and-fill

Raw materials are used for cut-and-fill first

EPS
construction method

Expanded polystyrene is not able to recycle, so the recycle is not consideredLightweight
embankment
construction method

Mixed breaking foamed styrol used once, soil andcement are used for lightweight embankment
construction method

After removing beads from dismantled banking
material, foamed styrol is recycled

Foamed waste glass
construction method

Raw materials of glass are used for foamed waste
glass construction method first

Dismantled banking material is recycled.

Fig. 3 System boundary of foamed waste glass construction method.
631

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 4. The CO2 emission amount results, fromlarger to smaller, are as follows: the EPS method,
the lightweight embankment construction methodwith EPS beads (without recycle and with recycle),
the foamed waste glass construction method(without recycle and with recycle), and the cut-andfill method. The EPS method results were due to thehigh amount of CO2 emitted by the manufacturing
and burning of EPS during its raw material
collection and construction stages. In the lightweight
embankment construction method with EPS beads,
most CO2 is emitted during construction anddisposal. This is due to the heavy equipment used in
the work. With respect to recycled materials, theCO2 generated during raw material collection is
reduced. The CO2 emitted by the foamed waste
glass construction method occurs mostly in theconstruction stage; hence, CO2 reduction is achievedby using recycled materials rather than by collecting
raw materials.
SOx emissions by the four construction methods,
from largest to smallest, are in the following order:
foamed waste glass construction method, foamed
waste glass construction method without recycle,
foamed waste glass construction method withrecycle, lightweight embankment construction
method with EPS beads without recycle, lightweight
embankment construction method with EPS beads
with recycle, cut-and-fill method, and the EPS
method. Large emission impacts were observed in
the raw materials acquisition and construction stages
in the foamed waste glass construction method, inthe construction stage alone in the lightweight
embankment construction method with EPS beads
and in the cut-and-fill method, and in the raw
materials collection and disposal stages in the EPS
method. SOx emission during raw materials
collection was reduced using recycled embankment
materials in the lightweight embankment
construction method with EPS beads and in the
foamed waste glass construction method.
The amounts of NOx emissions, from larger tosmaller, are as follows : the lightweight embankment
construction method with EPS beads (without recycle
and with recycle), the foamed waste glass
construction method (without recycle and withrecycle), the EPS method, and the cut-and-fill method.
Emissions during the construction stage had thegreatest impact in the lightweight embankment
construction method with EPS beads, the foamed
waste glass construction method, and the cut-and-fillmethod. In the EPS method, in contrast, emissions
during the raw materials collection and disposal
stages had greater impact. However, NOx emissions
during raw materials collection was reduced, as with
Sox, by using recycled embankment materials in thelightweight embankment construction method with
EPS beads and in the foamed waste glass construction
method.

database of JEMAI-LCA PRO [9].” The cost unit
for each material was estimated using the InputOutput Table of Japan’s Ministry of Internal Affairs
and Communications [10].
Table 3 CO2, SOx and NOx unit.
Life-cycle

Raw
Materials
Acquisition

Construction

Materials

Waste

EnergySOx

NOx(g-SOx)(g-NOx)

Soil, Sand (kg)0.00200.00340.0106

Limestone (kg)0.00470.00090.0015

Foamed EPS (kg)1.31230.2555

1.1651

Aluminum (kg)

9.218

76.8

30

Zinc (kg)

1.443

5.92

1.327

Porous lightweightfoam material

0.176

0.268

0.082

Leveling (m2)

20.90028.91948.17312.10016.74227.890

1.180

1.450

3.640

15t Truck (Diesel)
（km）

0.962

1.180

2.970

10t Truck (Diesel)
（km）

0.742

0.910

2.229

4t Truck (Diesel)
（km）

0.472

0.560

1.450

2t Truck (Diesel)
（km）

0.323

0.400

1.000

Dismantlement andrecycle (kg)0.001960.003410.01060

2Compaction (m )20t Truck (Diesel)（km）Transport

CO2(kg-CO2)

EPS (kg)

2.64

0.544

1.22

Metal (kg)

0.366

0.325

0.591

Diesel (L)

0.069

2.999

0.005

Electricity(Thermal power plant)
(kwh)

0.425

0.170

0.130

Table 4 Cost unit for each material.

Materials
Soil, Sand
Polystyrene
Limestone
Aluminum
Zinc
Additive agent
DieselElectricity
Leveling & Compaction

Unit
JPY/kg
JPY/kg
JPY/kg
JPY/kg
JPY/kg
JPY/m3
JPY/L
JPY/kwh
JPY/m2Cost
2.000
209.8620.633
76.539186.132
1000.000
78.00016.198934.271

RESULTS
Estimated amount of air pollutants and CO2
emissions of each construction method is shown in
632

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 5 shows estimated total LCCs of eachconstruction method. LCCs, from higher to lower, arein the following order: the lightweight embankment
construction method with EPS beads, the EPS method,
the foamed waste glass construction method, and the
cut-and-fill method. The raw materials collection anddisposal stages were more costly in the lightweight
embankment construction method with EPS beads
and in the cut-and-fill method. In the EPS and foamed
waste glass construction methods, more than 90% of
the costs occurred during the raw materials collection
stage. However, in the lightweight embankment

Fig. 4 Estimated amount of air pollutants and CO2 emissions of each construction method.

Fig. 5 Estimated total life-cycle cost of each construction method.
633

GEOMATE- Bangkok, Nov. 14-16, 2016

construction method with EPS beads and in the
foamed waste glass construction method, the cost of
raw materials collection was reduced by recycling
embankment materials.

health impacts must also be included in life cycle
impact assessments. In addition, it is necessary to
perform comprehensive evaluation that included theperspective of safety degree for a fair comparison
through life cycle impact assessment.

DISCUSSION
In this research, it was shown that all construction
methods need to recycle or reuse for lower negative
impacts because to produce new materials was bigger
adverse impacts than recycling or reusing materials.
As the evaluation results of CO2, air pollutantsand LCC, conventional cut and fill method was
lower impacts to environment and reasonable life
cycle cost than other methods. Regarding thelightweight embankment construction method, it
was bigger impacts on environment and cost in spite
of “with recycle” because CO2 and air pollutants atthe construction phase including manufacturing
process were greatly influenced in particular. Thus,
it was pointed out that it is necessary to improve thelower impact techniques to produce and recycle EPS
beads.
EPS method was the biggest CO2 emission thanother methods. We need to develop new technics torecycle or reuse EPS after dismantlement because
EPS blocks were just dumped now due to no recycle
and reuse method. On the other hand, Foamed waste
glass construction method with recycle was aroundsecond lowest impacts than other methods.
This research just only focused on CO2 emissions,
air pollutants and cost. Thus, we should selectappropriate construction method based on not only
those results but also those environmental
performance and regional characteristics onconstruction site.

REFERENCES[1] H. Ito, K. Yamanaka, S. Nehashi, A. Fukuda
and K. Minegishi, “Comparative Analysis ofThree Embankment Methods for Road
Construction by Life Cycle Assessment andCost”, International Journal of GEOMATE,
March 2015, Vol.8, No.2, pp. 1278-1282.
[2] H. Ochiai, K. Omine, “Evaluation ofenvironmental impact and development of
mixed composite materials with new additional
value.” The Foundation engineering &equipment, Vol. 32, Aug. 2004, pp.9-11.
[3] S. Inazumi, H. Otsu, T. Isoda, Y. Shigematsu“Evaluation of social environmental efficiency”.
on recycling of construction sludges as groundmaterials” Journal of geological engineering,
Vol.68. 2012, pp.163-174.
[4] K. Onitsuka, M.Yokoo, Y. Hara, S. Yoshitake,“Engineering Properties of Foaming Wasteglass Material and an Example of its
Utilization”, Journal of Soils & Foundations,Vol.47(4), 1999, pp.19-22.
[5] Architectural
InstituteofJapan,“LCA guideline for building”, 2006.
[6] Advanced Industrial Science and Technology,“IDEA (Inventory Database for Environmental
analysis)”, 2010.
[7] Life Cycle Assessment Society of Japan,“LCA database”, 2003.
[8] Express Highway Research Foundation of Japan,“Proposal of Estimation Method of CO 2
emissions on Road Projects”, 2004.
[9] Japan Environmental Management Association
for Industry, “JEMAI-LCA PRO”, 2000.
[10] MinistryofInternal
AffairsandCommunications, “Input-Output Tables ofJapan”, 2005.

CONCLUSION
In the present research, the potential for reducing
environmental load and LCCs using recycledmaterials in various embankment construction
methods has been proved.
For further research, estimated LCCs must
include external costs and considerations such as634

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

TSUNAMI IMPACT ANALYSIS TO GEOLOGICAL LANDSCAPE IN
PERAK COAST, MALAYSIA
1,2

Sukhveender Singh Sukhdev Singh1 and Abdul Halim Abdul Latiff2
Faculty of Geosciences and Petroleum Engineering, Universiti Teknologi PETRONAS, Malaysia

ABSTRACTOn the 26th December 2004, the world had witnessed one of the strongest ever seaquake measured at 9.1 M w
caused a highly catastrophic tsunami event, taking over more than 300,000 lives. The mega-earthquake wasbelieved due the rupture of a segment within the 1200 km fault line which lies off the coast of Sumatra, raising
the seafloor to over 8 meters in height, and produced the tsunami wave that was travelling at a speed of nearly
800 km/h. In this study, the presence of coastal vegetation and geological landscape were analyzed to reduce thetsunami impact, as well as determining the potential escape route in the event of the tsunami. In addition, atsunami computer simulation was performed along the Malacca Straits with reference to the earthquake source
located at Andaman Sea. The findings from this simulation were integrated with coastal profiling of Perak region
to produce a possible geological impact scenario and the potential evacuation plan, in the event of future stronger
earthquake. The resultant outcomes shall be shared with local government for a better tsunami disastermanagement procedure in Malaysia.
Keywords: Tsunami; Coastal Vegetation; West Malaysia Coastline
INTRODUCTIONBACKGROUND STUDYThe Sumatra-Andaman
earthquake that
occurred on the 26th December 2004 with themagnitude of 9.1 produced a tsunami that affected
the coastlines adjacent to the Indian Ocean, killed
more than 300,000 individuals, and brought aboutloss of property and vocation [1]. In definition,
tsunami is a gigantic wave that contains longwavelength and period which generated by thesudden, rough and impulsive undersea aggravations.
Based on the statistic, a total of 10 major tsunamis
have taken place around the world in between theyear of 2004 to 2013, which has taken lots of lives
with damages costing up to hundreds of billion
dollars [2]. Malaysia was one of the 13 countries in
the Indian Ocean that were flooded by the 26th
December 2004 mega-thrust earthquake. It wasrecorded that the total of 68 lives were taken when
the tsunami hit Malaysia, where the highest numberof casualties was in Penang (52 deaths), Kedah (12deaths), Perak (3 deaths), and Selangor (1 death) [3].

The study area chosen for this research work is
in Pasir Panjang beach, Segari, along the coast of
Perak, about 28 km north of Lumut. The Pasir
Panjang coastline is 8 km in length, 20 m in width,and almost in straight line with no particular barriers
blocking the giant waves from reach the beach (Fig.
1). The open and straight nature of the beach made it
highly vulnerable to the potential tsunami impact.
Our general observation found that, there is a slightelevation just by the coast of the beach with oil palm
trees were located few meters away from theshoreline. This vegetation can be acted as thepotential natural barriers to the tsunami with further
studies being carried out to prove this. There are alsoa few buildings such as Segari Turtle Care andBaitul-Hilal complex, which were constructed as a
tourist attraction zone.
Due to the lack of coastal study conducted inthis region, we took the initiative to look into thepotential tsunami impact on the coastal vegetationand geological landscape in Segari, west coast ofPeninsular Malaysia, by determining the tsunami
wave height run-up and its corresponding inundation
distance. The run-up which represents the verticallimit of the tsunami wave can be determined byconducting a computer simulation of the flooded
coastal area. The detailed study on the tsunami
inundation will determine how far tsunami can travel
onshore which might affect the coastal vegetationand geological landscape of the study area.

Previously, various studies and researches werecarried out with respect to the potential tsunami in
the west coast of Peninsular Malaysia, especially in
the Penang coastline where the beach was badly hit
by the tsunami. To complement those studies, thispaper will be focusing mainly in the coasts of Perakdue to the lack of study and research was previously
conducted in this area.

635

GEOMATE- Bangkok, Nov. 14-16, 2016enhance our understanding in potential catastrophic
tsunami. The resultant simulation model in term ofwave height than was analyzed and integrated withthe collected coastal data information.
COASTAL ANALYSIS
From the fieldwork conducted at the Pasir
Panjang beach, a beach profile map was developed.
The method used is the manual traversing along theshoreline and record all the surface information. It
was observed that the water depth away from theshoreline gradually become deeper with a gentle
slope, while the elevation of the beach is increasing
significantly inland (Fig. 2). Higher groundavailability in tsunami prone area is highly crucial in
reducing the number of casualties in the event oftsunami.
Another observation that could be noted was thesize of the sand grains. It was found that the finegrain sands filled the south-west part of the Pasir
Panjang beach. Traversing towards the north-eastside of the beach, the sand is getting coarser with amix of fine sand and eventually was just filled withthe coarse grained sand at the north of Pasir Panjang
beach.

Fig. 1 Location map of Pasir Panjang beach in
Perak coastline (indicate by the two pins).
METHODOLOGYAt the beginning of the study, a couple offieldworks were conducted to scrutinize few
important aspects related to the potential tsunami
impact such as:• Geological landscape and topography (e.g. beachprofile, ground elevation).
• Presence of coastal vegetation as natural barriers.
• Potential escape route.
All these aspects were observed and studied in orderto reduce the potential tsunami impact that mightoccur in this coastal area.
The information obtained then was used as an
input to tsunami computer simulation that conductedin the latter stage. The simulation method used tomodel the tsunami wave propagation that originatedfrom Andaman-Sumatra, is based on the Shallow
Water Equation (SWE). There are a number ofcontrols parameters that require close attention, withthe source and bathymetry data are the main
parameters that being experimented. Within thecontrols, there are a few parameters requiring input
in each of the domain.

Fig. 2 Coastal map of the Pasir Panjang beacharea where mix of fine and coarse sand can beobserved.
BATHYMETRY ANALYSIS
Before the tsunami simulation work can be
carried out, the bathymetry data which was obtained
from National
Oceanic
andAtmospheric
Administration (NOAA) need to be analyzed (Fig.
3). Since the tsunami wave originated fromAndaman Sea and propagated into the Strait ofMalacca, both bathymetry data were acquired fromNOAA as well as from Malaysian authorities. Thisbathymetry map then was used as a reference toproduce the tsunami wave simulation in determining
the wave run-up and inundation distance at the PerakThe source of tsunami wave generation can beinitiated by using either one of the two available
options; Gaussian hump in the initial tsunami
generation, or incorporating the Okada model togenerate the tsunami source [4]. Parameters such thecoordinate origin and the nature of the earthquakethat producing tsunami wave, were experimented inorder to initiate the Gaussian hump. Throughout thesimulation period, the dip, rake and focal depth ofthe earthquake are also being manipulated, to636

GEOMATE- Bangkok, Nov. 14-16, 2016

coastal region.

towards the study area. As the shallow waterequation has pointed out, the wave amplitude
indicates a significant decrease as the wave
approaching the shoreline (Fig. 5). However, ahigher precision wave simulation was not achievable
due to the difficulties of obtaining the suitablebathymetry format that suit the computer program.

Fig. 3 Bathymetry map of the Andaman Sea andStrait of Malacca with reference to the earthquake
and tsunami source in the Sumatra-Andaman
subduction zone.
TSUNAMI SIMULATION
Before embarking into tsunami simulation using
Shallow Water Equation (SWE) method, asimplified tsunami simulation was first carried out
by using interactive tsunami simulation program thatavailable online [5]. The program known as
Tsunami Mapper, use the flood fill algorithm withthe combination of elevation data provided in the
google map. By using Andaman subduction zone as
the point source (based on the 26th December 2004
mega-earthquake,) we set the maximum wave heightof 50 meters and the wave energy propagation
across the Indian Ocean of more than 1000 km.

Fig. 4 The potential tsunami wave propagationfrom the Sumatra-Andaman subduction zone
towards the Strait of Malacca, by using the Tsunami
Mapper program.

The resultant tsunami wave could be seen
spreading and propagating into the Strait of Malacca
where it will directly hits the west coast ofPeninsular Malaysia (Fig. 4). Tsunami wave wasalso predicted to reach the shore of Perak,
specifically in Pasir Panjang beach, where it is an
open coast and might impact the local reside at thearea. However, it should be noted that the simulation
provided by this program is highly inaccurate as
several factors was not taken into account; theearthquake characteristics such as magnitude, focal
depth and orientation, the bathymetry of the affectedarea and the wave energy at the shoreline. Thus, abetter and comprehensive program and algorithm isrequired to carry out further simulation in order todetermine the accurate wave run–up and inundation
distance at the coastal area.

Fig. 5 Tsunami wave propagation model of the
Pasir Panjang beach area.

To overcome the shortcoming shown in
Tsunami Mapper program, a simulation was carriedout by using TUNA-M2 software [6]. The software
which contains all the missing components that were
not available in the previous program, provide an
accurate tsunami wave’s direction and propagation

TSUNAMI IMPACT ANALYSIS
In general, the impact of the tsunami wave ismore severe if the shoreline topography lack of637

GEOMATE- Bangkok, Nov. 14-16, 2016

significant vegetation and elevation that can act as anatural barrier. To understand this impact, a crosssection map of the seafloor (Fig. 6) was made basedon the beach profile map generated in Fig. 2. Thiscross section map aids in the assumption on
determining the wave run up and inundation in orderto determine the potential impact tsunami wave hason the coastal vegetation and geological landscapeof the study area.

Nevertheless, data observation along the Pasir
Panjang beach indicates the previous 2004 tsunami
event had not produced the predicted wave run-up
and inundation distance simulation. To this extent,
we modify the earthquake’s fault orientation, from
Northwest-Southeast to North-South orientation [7].
By changing the fault orientation, the tsunami
impact from the Sumatra-Andaman earthquake ismore severe as the Perak coastal line is now facing
directly the tsunami wave propagation. The
maximum wave height simulated was able to reach 6
meters tall (Fig. 9).

Fig. 6 Cross section of the Pasir Panjang beachbathymetry and coastline.

Fig. 9 The projected maximum wave height at the
Pasir Panjang beach based on the modified
earthquake parameter.

thBased on the 26 December 2004 earthquake
characteristics, the simulation indicates that wave
height along Pasir Panjang beach able to reach 2.5
meters along the shoreline (Fig. 7). We used thisinformation to estimate the wave’s inundation
distance. Since there is ground elevation by thecoastal area, the simulated inundation distanceindicates that the water could flood up to 55 metersfrom the shoreline (Fig. 8).

In the previous simulation, it was found that, thesimulated inundation distance along the Strait ofMalacca is generally around 15 to 20 times longer
than the recorded wave height [7]. Due to theshallow bathymetry and gentle coastal gradient, thisassumption is valid for the west coast of Peninsular
Malaysia and Strait of Malacca. It can be calculated
that the corresponding tsunami wave’s inundation
distance will be around 90 meters (Fig. 10).
However, it should be noted that several factors suchas wave velocity and gravity were not taken inconsideration in producing this model, thus making
this inundation model to be slightly inaccurate.
Nonetheless, this model outcome shall be used as areference or guidance to aid tsunami education andawareness among the Pasir Panjang’s coastal
community.

Fig. 7 The maximum wave height at the Pasir
Panjang beach based on 26th December 2004
earthquake and tsunami characteristics.

Fig. 10 The projected maximum inundation
distance at the Pasir Panjang beach based on
modified earthquake parameter.

Fig. 8 The maximum inundation distance at the
Pasir Panjang beach based on December 26, 2004
earthquake and tsunami characteristics.
638

GEOMATE- Bangkok, Nov. 14-16, 2016

to its shoreline. From the analysis of both theprevious 2004 tsunami event and the possible
maximum scenario tsunami event which hasdifferent earthquake fault orientation (Table 1), it
can be deduced that an emergency evacuation routeshould be determined. In addition, the locals should
be briefed regularly about tsunami evacuation
process in order to prepare for any possibilities, eventhough a tsunami of this magnitude may notoccurred for the nearest time. It also suggested that awave breaker should be built near to the coastline tominimize the casualty in the event of tsunami.

EVACUATION ROUTE RECOMMENDATIONThe chalets, turtle sanctuary and other buildings
just close by the beach will be affected if a bigger
tsunami wave hit the study area. However, thecoastal elevation of around 1-1.5 meters might
reduce the strength and impact of the incoming wave.
A few oil palm plantations will also be affected inthis event but will help in as a “natural barrier” toprevent the tsunami wave from further travellinginland. In this particular case, it would be wise toevacuate the people around this tourist andrecreation area earlier and head to the nearest tallest
building that is located right behind the Segari Turtle
Sanctuary (Fig. 11). It should be note that there are
only a single lane and narrow road into and out fromthe Pasir Panjang beach area. Thus in the event of a
massive tsunami wave during a festive season, theroad will become highly congested, preventing anefficient evacuation process. Therefore, it is also
recommended the local authority to widen the roadand built an alternative road lead to and out fromthis popular tourist spot.

Table 1: Comparison of 2004 tsunami simulation to
maximum scenario tsunami simulation.
Parameter
Magnitude
Focal Depth
Fault Orientation
Wave HeightRun-up Height
Inundation Distance

2004
Tsunami
9.0
30 km
NW-SE
~2.5 m~3.5 m35 – 45 m

Maximum
Scenario
9.0
30 km
N-S
~5.5 m
~6.5 m80-90 mACKNOWLEDGEMENT
We would like to thank Universiti Teknologi
PETRONAS for funding this research project.
REFERENCES[1] Colbourne, F. W. (2005). Tsunami Impact on
the West Coast of Penang Island, Malaysia.
Emporia State University.
[2] National
Oceanic
andAtmospheric
Administration. Retrieved October 13, 2015(http://www.tsunami.noaa.gov/tsunami_story)[3] Penang Office for Human Development (2005).
Retrieved
December
18,
2015(http://www.pohd.org/media/Files/pgreport)[4] Okada, Y., (1985). Surface Deformation due to
Shear and Tensile Faults in a Half-Space,
Bulletin of the Seismological Society ofAmerica, Vol. 75, No. 4, pp. 1135-1154.
[5] Tsunami Mapper. Retrieve June 9, 2016(http://tsunami-maps.com).
[6] Koh, H. L., & Teh, S. Y., (2010). TUNA User
Manual.
[7] Koh, H. L., Teh, S. Y., Majid, T., Lau, T. L., &Ahmad, F. (2011). Earthquake and Tsunami
Research in USM: The Role of Disaster
Research Nexus.

Fig. 11 Potential evacuation route in the event oftsunami in Pasir Panjang beach area. Safe zone
indicate the high-rise building that can be used as the
shelter.

CONCLUSION
In a nutshell, the general idea of the tsunami
impact towards Perak coastal area, particularly inPasir Panjang beach has been presented from thisstudy. It is clear that Pasir Panjang beach in the west
coast of Peninsular Malaysia, is an area that highly
vulnerable to tsunami as it is located along the Straitof Malacca and without any coastal protection near639

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

THE EXPERIMENTAL DESIGN AND CARBON FOOTPRINTASSESSMENT OF NON-GLAZED FLOOR TILES
Karin Kandananond
Faculty of Industrial Technology, Valaya Alongkorn Rajabhat University, Thailand

ABSTRACT
Carbon emission from the manufacturing sector is a critical issue which is concerned by the environmental
authorities since the violation of the carbon emission cap might lead to the sanction by one of Thailand's largest
trade partner, European Union (EU). As a result, it is important for the manufacturers to be able to assess theirown products' carbon footprint. In this study, the selected case study is a ceramic factory which manufactures
non-glazed floor files. The scope of evaluation covers Business-to-Customer (B2C) transaction while the life
cycle of a product includes four stages, i.e., resource extraction, manufacturing, distribution, use and waste
disposal. The study results indicate that the highest contribution to the carbon emission is from the extraction ofceramic clay while the manufacturing stage has the second highest effect on the emission. The distribution of
products, use and disposal are the life cycle stages which have small effects on the emission. Another objectiveof this research is to conduct an empirical study which leads to the capability to quantify the effect of different
factors on the manufacturing of floor tiles. According to the experimental study, three factors, i.e., chalk clay,
ball clay and feldspar, are considered as the process inputs while the response variables are percent absorption
and hardness. Elaborately, 23 full factorial design was deployed to study the find the relationship between inputs
and outputs. The results has two folds. The first fold is useful for the manufacturers who would like tounderstand how much their product has emitted the greenhouse gas to the atmosphere and it might lead to theminimization of their emission. Moreover, the relation between the tile characteristic and factors affecting the
manufacturing is known so the manufacturer is able to efficiently optimize the manufacturing process in order toachieve the highest quality products.

Keywords: Business-to-Customer, Carbon footprint, Design of experiment, Life cycle, Non-glazed floor tile.

average emission per functional unit to theatmosphere.
Among many industries, ceramic manufacturing
is among the industries which are responsible for theemission of a large amount of GHG in Thailand.
According to the report by the European ceramicindustry association, the emission due to the ceramic
production will be reduced if all the kilns used in theindustry are improved to fire products efficiently [1].
In 1998, the National Pollutant Inventory unit of the
Queensland department of environment issued theemission report for bricks, ceramics, clay andproduct manufacturing [2]. Quinteiro, Araujo, Dias,
Oliveira and Arroja [3] conducted a study tocompute the GHG emission of different ceramicearthernware pieces. Similary, Quinterio, Almeda,
Dias, Araujo and Arroja [4] had extended theirresearch in 2012 to cover other ceramic products, i.e.,
brick, roof tile, wall, floor tile and sanitary ware.
Peng, Zhao, Jiao, Zheng and Zeng [5] havecalculated the CO 2 emission and also suggest theoptions to reduce the emission in a ceramic tile
manufacturer. For construction purpose, Sazedj,
Morais and Jalali had compared the CO 2 emission
from two types of materials, bricks and concrete

INTRODUCTION
Greenhouse gas (GHG) emission is the criticalissue which is on the spotlight of the worldcommunity since it is one of the possible causes of
the world’s climate change. There are initiations
from developed countries to address the issue andalso the resolution to reduce the emission. European
Union (EU) is among the very first group of nations
which create the awareness by issuing the carbon
credit and carbon footprint schemes. On the other
hand, President Barack Obama of the United States
has declared in 2015 that the US will decrease theemission for one-third by reducing the emission
from power generation which is based on coalburning power plants. For the emerging economies
like Southeast Asian countries, Thailand is one of
the leading nations which is aware of the carbon
footprint issue since it has the official organizationbody (Thailand Greenhouse Gas Management
Organization or TGO) which is responsible formanaging the GHG emission in Thailand. As a result,the manufacturers in Thailand have the guidelines
for calculating the carbon footprint of their products
and the assessment will lead to the awareness of the640GEOMATE- Bangkok, Nov. 14-16, 2016

block [6]. Bribian, Capilla and Uson [7] also studied
the energy demand and CO 2 emission among
different construction materials which are ceramic,
steel, PVC, wood, mortar, cement, aluminium andlime. The gas release during firing of clay to
produce bricks was studied by Toledo, Santos, Faria,
Carrio, Auler and Vargas [8]. In this study, theexperiment was conducted to examine the amount ofemission at the different temperatures.
According to the literature, the emission from theproduction of different ceramic products is studied
while the focus is on the construction materials.
However, the emission from small manufacturers
seems to be ignored even it also contributes a large
portion on the emission since, in Thailand, mostceramic manufacturers are small and medium
enterprises. The awareness regarding the emission isimportant to both manufactures and consumers. The
selected case study for the emission assessment inthis research is the emission from the whole life
cycle of a ceramic product, non-glazed floor tiles
manufactured in a small factory. Elaborately, theemission from each stage of the life cycle isprofoundly analyzed and calculated. Moreover, thestudy also points out the hotspot which is highly
contributed to the major emission. Last but not least,
the recommendation for the reduction in theemission is also addressed and discussed.

Fig. 1 Sample of floor tiles.

MANUFACUTRING PROCESS
Fig. 2 Excavation site.
According to the study, the instructions by
Thailand greenhouse gas management organization(TGO) are carefully followed while the evaluation isbased on the transaction of B2C (businesss-tocustomer). A product chosen as a case study in thisresearch is a non-glazed floor tile as shown in Fig. 1.
The weight of a floor tile (functional unit) is 0.225
kg. The main ingredient of the floor tile is the clay
excavated from the area of Lampang province (Fig.
2 and 3). Other ingredients are Feldspar (20% of thefloor tile weight = 0.045 kg) and Kaolin (20% of thefloor tile weight = 0.045 kg). The production facility
is located in Prathumthani province. The kiln used
for firing tiles is illustrated in Fig. 4 and the sourceof fuel for this kiln is liquefied petroleum gas (LPG).
Another important aspect for the life cycle analysis
is the clarification of manufacturing process. For
producing floor tiles, the manufacturing process
consists of the following steps, forming andfinishing, biscuit firing and glost firing and it isconcluded in a flow chart (Fig. 5).

Fig. 3 Excavation.

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Table 1 GHG emission of raw materials extraction
Raw
Materials

Quantity(kg)

EF(kgCO 2
eq/kg)

Clay
FeldsparKaolin
Total

0.135
0.045
0.045

0.86350.2167

EmittedGHG(kgCO 2eq)
0.03886
0.0097515
0.0486115

Another important source of emission whichcannot be ignored is the emission due to thetransportation of raw materials to the factory. Afterthe excavation, the clay is shipped to the factory
which is located in Prathumthani province by a tenwheeled truck with the maximum load of 16 ton.
The distance between the factory and the source ofclay is 568 km. On the other hand, other ingredients(Feldspar and Kaolin) are transported from a
supplier in Bangkok by a four-wheeled truck withthe maximum load of 75% of 7 ton. The supplier’s
warehouse is 65 km from the factory. The
computation is separated into two parts, delivery andreturn. In Table 2, the load of the transportation is
illustrated in the form of tkm unit. The GHG
emission of the transportation of raw materials forthe delivery trip is shown in Table 3.

Fig. 4 Kiln fuelled by LPG.

Table 2 Load of raw material (delivery)
Raw
Material
Clay
FeldsparKaolinQuantity
(kg)
0.135
0.045
0.045

Distance
(km)
568
65
65

Load
(tkm)
0.07668
0.002925
0.002925

Table 3 GHG emission of the transportation of raw
materials (delivery)
Fig. 5

Manufacturing process.

LIFE CYCLE ANALYSISThe life cycle analysis of non-glazed floor tile isdifferentiated into five stages, resource extraction,
manufacturing, distribution, use and waste disposal.

Raw
Material

Mean ofTransportation

Clay

Ten-wheeled
Truck (max load:
16 ton, 100%
Loading)
Four-wheeled
Truck (max load:
7 ton, 75%
Loading)
Four-wheeled
Truck (max load:
7 ton, 75%
Loading)

FeldsparKaolinResource Extraction
The main ingredient of the floor tile in thisresearch is the clay excavated from a paddy rice
field which is located in Lampang province. Other
two ingredients are Feldspar and Kaolin and theiremission analysis is shown in Table 1 which
illustrates the quantity in weight per unit andemission for each ingredient.

Total

EF
(kgCO 2
eq/tkm)0.0451

EmittedGHG(kgCO 2 eq)
0.003458268

0.239

0.000699075

0.239

0.000699075

0.004856418

Besides the delivery of raw materials, the emission
from the return trip of the trucks has to be includedin the calculation as shown in Table 4 and 5. For the
return trip, the maximum load of the mean oftransportation must be included in the calculation asrecommended by TGO. For example, the load ofclay is equal to 0.135*568/(1000*16) = 0.0047925
642

GEOMATE- Bangkok, Nov. 14-16, 2016

LPG use is for two processes, biscuit firing and glost
firing. The total emission due to the manufacturing
is shown in Table 8.

tkm.
Table 4 Load of raw material (return)Raw
Material
Clay
FeldsparKaolinQuantity
(kg)
0.135
0.045
0.045

Distance
(km)
568
65
65

Load
(tkm)
0.0047925
0.000417857
0.000417857

Table 8 GHG emission of the manufacturing (LPG
use)
Resouces

Process

Quantity(kg)

EF(kgCO 2
eq/kg)

LPG

BiscuitFiring0.2

0.4116+
0.4122LPG

GlostFiring0.3

0.4116+
0.4122

Table 5 GHG emission of the transportation of raw
materials (return)Raw
Material

Mean ofTransportation

Clay

Ten-wheeled
Truck (max load:
16 ton, 0%
Loading)
Four-wheeled
Truck (max load:
7 ton, 0%
Loading)
Four-wheeled
Truck (max load:
7 ton, 0%
Loading)

FeldsparKaolinEF(kgCO 2
eq/tkm)0.5711

EmittedGHG(kgCO 2 eq)
0.002736996

0.33240.000138895

0.33240.000138895

Total

Total

Table 6 GHG emission of the transportation of raw
materials

Extraction
Clay
FeldsparKaolin0.03886
0.0097515

Table 9 Weight and distance per unit of floor tile(delivery)
Product

Transport
Return0.002736996
0.000138895
0.000138895

Floor tile

Clay
FeldsparKaolinTotal

Weight(kg)
0.225(1Piece)

Distance
(km)
75

Load
(tkm)
0.016875

Table 10 GHG emission due to the delivery per unitof floor tile

Table 7 Total GHG emission of the transportation of
raw materials
Raw Material

0.4119Since the distribution stage solely relies on thetransportation, the computation of GHG emission issimilar to the shipment of raw materials to thefactory. After the manufacturing stage, floor tiles as
the finished product are shipped to the Chatuchak
Sunday market which is 75 km away from thefactory. For the delivery trip, the emitted GHG iscalculated as shown in Table 9 and 10 respectively.

The total emission of the raw materials regarding thedelivery and return trip is concluded in the following
Table 6 and 7.

EmittedGHG(kgCO 2 eq)
Transport
Delivery
0.003458268
0.000699075
0.000699075

0.24714

Distribution

0.003014786

Raw
Material

EmittedGHG(kgCO 2eq)
0.16476

Total Emission(kgCO 2 )0.006195264
0.03969797
0.01058947
0.056482704

Product

Mean ofTransportation

Floor tile

Four-wheeled
Truck (max load:
7 ton, 100%
Loading)

EF(kgCO 2
eq/tkm)0.1402

Emitted GHG(kgCO 2 eq)
0.002365875

On the other hand, the return trip for a truck has
emitted the amount of GHG gas as shown in thefollowing Table 11 and 12 while the total emission
is depicted in Table 13.

Manufacturing
The manufacturing process basically depends onfiring floor tiles and it is expected to be the main
source of GHG emission. The fuel used in the kiln isliquefied Propane gas (LPG). The emission due to
LPG is divided into two substages, acquisition anduse. The emission factors for acquisition and use are
0.4116 and 0.4122 kgCO 2 eq/kg respectively. The

Table 11 Weight and distance per unit of floor tile
(return)
Product
Floor tile643

Weight(kg)
0.225(1Piece)

Distance
(km)
75

Load
(tkm)
0.0024107143

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 12 GHG emission due to the return trip perunit of floor tile
Product

Mean ofTransportation

Floor tile

Four-wheeled
Truck (max load:
7 ton, 100%
Loading)

EF(kgCO 2
eq/tkm)0.3111

Emitted GHG(kgCO 2 eq)
0.0007499732

Table 13 Total GHG emission regarding thetransportation
Product

Floor tile

EmittedGHG(kgCO 2 eq)
Transport
Transport
Delivery
Return0.002365875
0.0007499732

Fig. 6 Elaborated GHG emission by each category.

Total Emission(kgCO 2 eq)

0.0031158482

Use
Approximately, floor tiles are cleaned monthly toremove dust, mud and fungi. Water spray is used toclean the surface of floor tiles. The amount ofemission regarding the use is shown in Table 14.

Fig. 7 Overall GHG emission by category.
The transportation of raw material to the
manufacturer and the shipment of finished goods tothe market also rely on the fossil fuel. Therefore, if
there is a reduction in the use of fossil fuel, it will
significantly decrease the GHG emission to theatmosphere. On the other hand, if the alternative
source of energy besides fossil fuel with low GHG
emission is explored and used, it will lead to the
reduction in the carbon footprint of the wholeprocess as well.

Table 14 GHG emission due to the use
Resource

Quantity
(kg)

Water

0.2

EF
(kgCO 2
eq/unit)0.0003

Emitted GHG(kgCO 2 eq)
0.0006

Waste Disposal
Since this type of floor tiles is non-glazed and no
additional color is applied to the tiles (earth tone),the recycling rate is almost 100 percent. Therefore,
there is no GHG emission in this stage.

DESIGN OF EXPERIMENTThe 2k factorial design of experiment is deployed
to study the effect of three factors’ ingredients, ballclay (A), Kaolin (B), Feldspar (C) on two crucial
characteristics of tiles, namely, rate of water
absorption and hardness. Table 15 illustrating input
factors and their levels are shown as follows:

Overall emission
Due to the life cycle analysis, the total emissionof each stage is shown in Fig. 6 and 7 consecutively.
Most of the emissions fall into the category of
resource extraction and manufacturing process. The
implication is that the large portion of contribution
on the emission is caused by the raw material
supplier and manufacturer. Since the manufacturing
process depends on the firing by using LPG, theacquisition and use of fossil fuels is the main sourceof emission in this case.

Table 15 Total GHG emission regarding thetransportation
FactorBall Clay(A)Kaolin (B)Feldspar (C)644High (1)
10%
5%
10%

Low (-1)
30%
20%
20%

GEOMATE- Bangkok, Nov. 14-16, 2016

Design-Expert® Software

characteristic of floor tiles. For the life cycle
analysis, the manufacturing fraction seems to mostly
contribute on the greenhouse gas emission.
Therefore, the process improvement should focus on
the reduction of emission from the activities of the
manufacturing process.
According to theexperimental analysis of the desired characteristic,
the result shows that there are interactions among
two and three factors which affects the rate of water
absorption and hardness of floor tiles.

Cube
Absorb

Absorb
X1 = A: A
X2 = B: B
X3 = C: C

28

24

29

B: B

B+: 1.00

27

30

28

C+: 1.00

C: C

B-: -1.00
26
A-: -1.00

A: A

C-: -1.00
24
A+: 1.00

REFERENCES

Fig. 8 Cube plot for rate of absorption rate.

[1] European
ceramic
industry
association,“National Pollutant, “Emissions estimation
technique manual for bricks, ceramics, & clay”
1998.
[2] European integrated pollution prevention andcontrol Bureau, “Ceramic Manufacturing
Industry”, 2007.
[3] Quinteiro P, Araújo B, Oliveira B, Dias AC,
Arroja L, “The carbon footprint and energy
consumption of a commercially produced
earthenware ceramic piece”, Journal of theEuropean Ceramic Society, Vol. 32(10), 2012,
pp. 2087–2094.
[4] Quinteiro A, Almeida M, Dias AC, Araújo A,
Arroja L, “The Carbon footprint of ceramic
products”, EcoProduction, Vol. 1, 2014, pp.
113-150.
[5] Peng A, Zhao Y, Jiao L, Zheng W, Zeng L,“ CO 2 emission calculation and reduction
options in ceramic tile manufacture-the Foshancase”, Energy Procedia, Vol. 16, 2012, pp. 467476.
[6] Sazedj S, Morais AJ, Jalali S, “Comparison ofembodied energy and carbon dioxide emissionsof brick and concrete based on functional units.
Congresso Internacional, Luso-Brasileiro deMaterials de Contrucao Sustentaveis, Vol. 1,
2014, pp. 81–88.
[7] Bribian IZ, Capilla AV, Uson AA, “Life Cycle
Assessment of Building Materials: Comparative
Analysis of Energy and Environmental Impacts
and Evaluation of the Eco-Efficiency
Improvement
Potential”,
Building
andEnvironment, 2011.
[8] Toledo R, Dos Santos DR, Faria Jr. RT, Carrio
JG, Auler LT, Vargas H, “Gas Release During
Clay Firing and Evolution of Ceramic
Properties”, Applied Clay Science, Vol. 27,
2004, pp. 151–157.

If the interested characteristic is the rate of water
absorption, the statistical analysis indicates that there
is a three factor interaction (ABC) among all threefactors (ingredients), Ball Clay (A), Kaolin (B) andFeldspar (C). According to the cube plot in Fig. 8 ,floor tile will have the low water absorption rate
when the percentage ingredients of each factor are
set at the following levels:-A (low), B (high), C (low) and-A (high), B (low), C (low).
For the hardness, the result obviously shows that
there is an interaction between Kaolin (B) andFeldspar (C). The important finding is thepercentage of Kaolin is high (20%), it will result in
the reduction of the hardness of the floor tile nomatter the Feldspar component is set at the low orhigh level.
Design-Expert® Software

Interaction
C: C

Hardness452

C- -1.000
C+ 1.000
446.5

Actual Factor
A: A = 0.00

Hardness

X1 = B: B
X2 = C: C

441

435.5

430

-1.00

-0.50

0.00

0.50

1.00

B: B

Fig. 7

Interaction plot for rate of water absorption.

CONCLUSION
This objective of this research has two folds,
the life cycle analysis and the conduction ofexperimental analysis which leads to the desired

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Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

DAMAGE QUANTIFICATION OF BEAMS USING FREQUENCYSIGNATURE
1Bryan Josef T. Medrano1 and Lessandro Estelito O. Garciano, D. Eng2
Graduate student, De la Salle University - Manila, Philippines; 2 Associate Professor, De la Salle University- Manila, Philippines

ABSTRACTA rapid method for determining the damage severity sustained by a beam proved to be challenging due toeither limited studies conducted on the subject or alternative methods require highly sophisticated and costly
equipment to perform. In this research, the unique frequency signature emitted by a beam when excited by an
external force was utilized in order to determine the changes in the properties of the beam. Experiments were
performed using a roving accelerometer hammer impact test on a beam with a grounded configuration to test thechanges occurring as the controlled damage sustained by the beam increases. The acceleration response of thebeam obtained from the experiment is then processed using software incorporating Kalman Filter and structuraldynamics. Results show that the dominant frequency obtained in both the Fast Fourier Transform and PowerSpectral Density of the acceleration response of the beam decreases as the damage incurred by the beamincreases. The results also show that regardless of the position of the accelerometer, dominant frequencies tend
to converge to a value depending on the damage sustained in the beam.
Damping ratio of the beam also decreased as the damage sustained by the beam increased. Inversely, theincrease in damage of the beam corresponds to an increase in the dissipation rate of the beam. The study wasable to achieve its goal of quantifying damage in a beam through the use of frequency signature by identifying
the changes in its dominant frequencies and the damping ratio and dissipation rate.
Keywords: Damage Quantification, Impact Hammer Test, Frequency Signature, Damping Ratio
damage type, quantification of damage severity andprediction of the remaining life service of thestructure [2].
Several studies have been made in order todetermine the damage sustained by beams. A study
using the Artificial Neural Networks that makes useof global and local vibration-based analysis data as
input were conducted to determine the location andquantified depth of damage in beam like structures
were done in previous years [4]. Another recent
study on the other hand, makes use of guided waves
and Bayesian statistical framework for thecharacterization of the damage in beams [5]. Other
studies conducted made use of modal power flow
generated by structural elements subjected tovibrations in determining the crack length and depth[6], [7]. The most recent studies made for damage
identification made use of the acoustic emission
techniques which make use of state-of-the-art
equipment to determine the location and severity ofthe damage [8], [9].
This research therefore aims to provide a meansof detecting the level of damage severity a structuralelement has sustained. With this method, damage
can be quantified and the damages incurred by the
structure can be prioritized according to the degreeof severity. The need for the prioritizing the damage
sustained by the beams would be useful to further
understand the situation and be able to quickly
remediate the problem and avoid serious damage.

INTRODUCTION
In a modern community, engineering structures
play a vital role and therefore they are usuallydesigned to have long life spans. Human lives andproperty may be affected or disrupted by the failure
or unsatisfactory performance of these structures andis therefore necessary to ensure the safety andreliability of structural members.
Structures are prone to damage due to loadingfrom continuous use or due to stress caused byexternal loads such as earthquakes. With this,structural damage detection is a vital factor inmaintaining the integrity of the structure and reduces
the likelihood of structural failures which can have
grave consequences [1]. At present, the use of nondestructive examination is being used by engineers
in determining the damages of a structure without
the need to demolish the structural element of thebuilding being analyzed [2].
The method of detecting damage in a structure
should also consider the quickness of identifying the
problem in the structure. Vital structures including
hospital, bridges and fire stations require rapid
identification of damage in the structure in order toprevent secondary damage [3]. A recent study showsthat there are five levels of damage detection to beconsidered in monitoring the condition of a structure.
These are mainly the identification of existingdamage, localization of damage, identification of646

GEOMATE- Bangkok, Nov. 14-16, 2016

The repairs necessary for the structure can be
determined based on the output of the study.
Damage severity is one of the five key levels indamage detection of a structure [2]. Damage severity
is a necessary procedure when conducting damage
detection in a structural element of a building.
However, the lack of a quick and convenient methodfor determining the severity of a structural damage iscurrently not present. On the other hand, methods
for determining the damage severity of the structure
require serious amounts of data to provide accurate
results. There is a need to study a method for
determining and quantifying the severity of damage
in a structural element of a building.
As seen in several stated literatures, the main
focus of other studies is on damage identificationand localization. Damage quantification alone was
not dwelled in most literature. On the other hand,damage quantification that is determined along withthe other levels of damage quantification requires
state-of-the art machinery or tools to determine theunknown parameter.
Damage severity detection is crucial indetermining the status of a structure. It should bedetermined along with the other levels of damage
detection such as existence detection, localization ofdamage, damage type identification and predictionof remaining service life [2]. Fewer studies havebeen conducted to determine the severity of thedamages as compared to studies which determine thelocation and existence of damage. This gives all themore reason to conduct the study. Also, mostimportant structures are found in urbanized anddensely populated areas, the study is vital to ensure
the safety of both the structure and the people
occupying it.

The data is then analyzed to obtain the final output
in terms of a frequency of the fast Fourier transformand power spectral density as well as the damping
ratio of the element.
THEORETICAL FRAMEWORKThe study aims to identify the changes to the
dynamic properties of the box beam using thedifferent theories and techniques presented inprevious researches and published material. The
approach to the study involves the use of threetheories, i.e., Kalman Filter, direct integration
method and the approximation of the damping ratio.
Kalman Filter
The Kalman Filter is an optimal solution tofiltering problems which have systems andobservation models that are both linear and have
Gaussian probability density functions. With thefollowing assumption considered, the equations
taken from the optimal Bayesian equations would be
reduced to the following.
x k  fx k 1  w k

Eq. 1

z k  hx k  v k

Eq. 2

And Kk is the Kalman gain given asK k  f k 1 / k Pn / n1 hkT hk Pn / n1 hkT  Rk

,
1

Eq. 3

And Pk is the Variance-covariance matrix which
is given as

Pk  Pn / n1  f k / k 1 K k hk Pn / n1

CONCEPTUAL FRAMEWORKStructural elements gave off a frequency
signature when the element experienced an externalload. These frequency signatures were captured
through the use of a sensor to gather data and store
on to a laptop. Each beam gave off a uniquefrequency signature due to the minute differences
caused by the imperfections found in the beams.
Given that the structural element was set at aconstant properties and imperfections are set to aminimum, the frequency signature produced was
also presumed to be at a constant.
In order to test and determine the degree ofseverity that a structure has sustained, the frequencyof the beam was tested at different depth of damage.
The cracks or fractures of the beam was represented
by a chipped off portion and these damages are
varying in depth to record the change in frequency
signature. The recorded frequency signatures of thedamaged and undamaged beam were then compiled
and were then used in the Kalman filter program.

Eq. 4

Equation of motion for a discretized beamConsidering a finite element beam model
wherein the properties are distributed to eachelement, a motion of equation can be setup for eachelement present in the beam. Figure 1 shows atypical beam with an n-th number of elements. Eachelement is considered for analysis and that the
motion of the entire beam can be determined withrespect to the movement of each element. Similar toa single oscillating object, the elements of the beam
have properties such as mass, damping coefficient,
stiffness and external force acting on the element
which are independent from one element to the otherand values such as acceleration, velocity anddisplacement which are the basis for the motion of
the beam.

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GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 1

Typical Beam with n number of elements

Fig. 2

Free Body Diagram of a beam with an n-th
number of elements

Eq. 8
WhereinState of the system equation and Observation
equation
As one of the methods for direct integration of
second order derivatives, Newmark’s Constant
Average Acceleration method make use of the factthat the acceleration of a given time interval is equal
to a constant value. The algorithm numerically
updates the response acceleration, velocity anddisplacement of an object from t i to t i 1 . Following

Eq. 9
Considering the latter of the two equations, theobservation equation is based on the independent
variable that has been measured in the experimental
portion of the study. Given that the input actuatorsor the accelerometers each measures only theacceleration response of the material being tested,
the observation equation would be a vector matrix
and an identity matrix defined as

the previous stated assumption and isolating , theequation of motion becomes
Eq. 5

Using the truncated Taylor’s series expansion,
the following equation that would complement theprevious equation and complete the algorithm are
Eq. 10Eq. 6

Measuring Damping Ratio
In evaluating the damping ratio of a freevibrating structure, the logarithmic decrement
method can be used. The logarithmic decrement
method is used to measure damping in time domain.
In this method, the free vibration displacement
amplitude history of a system to an impulse ismeasured and recorded. Logarithmic decrement is
the natural logarithmic value of the ratio of two
adjacent peak values of displacement in free decay
vibration. Given that the damping ratio is small andwith the exponential in the ratio
can beexpanded in series retaining only the first two terms,
since
this leads to

Eq. 7
Incorporating previous equations and the new
equation becomes the state equation of the Kalmanfilter updates the acceleration based on the
numerical integration of the equation of motion andwould be used to compared the measured
acceleration obtain from the experimental portion of
the study.

Eq. 11
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GEOMATE- Bangkok, Nov. 14-16, 2016

dimensions of the damage.
For cases where the difference between two
amplitude peaks are very small, it is moreconvenient to choose two non-consecutive peaks andthe equation would be
Fig. 4 Dimensions of damage (a) Elevation (b) top
All damages on the beam trial were located on
the first quarter mark which was equal to 0.5m fromorigin.
The beams were supported on box filled withsand in order to make use of the boundary effects
when vibrations passed through the beam.
Once the pieces were sawed, it is then nailed andwood glue is placed on its corners to avoid breaking
from the weight of both the beam and the sand. A
hole with the exact size of the beam was thenpunctured and filled with sand up to the bottom
portion of the hole.
The input actuator with negligible mass that wasused for recording and measuring the response of thebeam was a triaxial accelerometer as seen in Fig. 5.

Eq. 12RESEARCH METHODOLOGY
Pre-experimentation phase
As the initial step to the study, the research was
required to design the appropriate beam in order toproduce result which will represent beams in its true
scale and purpose in a structure.
The following were considered in the design ofthe beam element which would represent thestructural element of a building and would provide
the necessary data needed for the study.
The beam element was made of a rectangularaluminum box beam, specifically the material was a6061 T6 3x3” Aluminum Square Tube.
As for the length of the beams, the original piece
used for the test were 3 pieces of standard length
which amounts to a total length of 21 ft. per piece orapproximately equal to 6.4m per piece. The eachmaterial was cut into 3 equal part wherein the 2m of
the beam was considered as the effective length of
the beam. The remaining portions of the beam wasburied inside the wooden box frame to dissipate
vibration that would reflect back. Figure 3 illustrates
the dimensions of the beam from the front view.

Fig. 3Fig. 5

Triaxial Accelerometer

The hammer that was for all trial in theexperiments was an ordinary rubberized hammer
typically used for basic construction and carpentry
Experimentation proper
With all beams cut according to specification, thebeam ends were covered using packaging tape toreduce the amount of unnecessary matter such as
sand from entering the beam.
Proceeding to the next step of the experiment,
five accelerometer were used for the undamaged
beam and six accelerometers for all damaged beams.
Given that the setup is ready and the triaxial
accelerometers’ capabilities were tested, thevibration test could now proceed. The vibration test
used for the experiments was the impact hammer
test. The design of the experiment incorporated theuse of a rubberized hammer that was dropped at thecenter of the beam which provided the maximum
movement possible for the beam to produce.
In Summary, Fig. 7 shows the complete
experimental setup

Box beam dimensions

The beams were classified in to two cases,
wherein there will be one undamaged and fivedamaged beam. The damaged beam were therepresentation of the deterioration or cracks
developed by the beam with each case having a
different degree of damage that was predeterminedin order to consider only the quantification ofdamage. The damage ranges from 1 to 4 percent of
the total length of the beam. Its design also had aconstant depth of damage equal to 50 percent of thetotal height of the beam. Figure 4 shows the649

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 7

The trend indicates that as the damage sustainedby the beam or any structural element increase, thefrequency produced by the said beam or element
decreases accordingly. Also, it could be pointed outthat the drop in frequency is consistent regardless ofthe position of the accelerometer whether it isposition nearest to the sensor or it is place nearest to
the point of impact or even positioned at the father
point from the damage. Table 1 established thepercent difference between the frequencies recordedby each sensors and the mean frequency of eachbeam case.

Complete experimental setup

Post-experimentation phase
Using the obtained acceleration data from theexperiments, the Kalman filter processed this data
for the observed stated and has been incorporated by
the use of the observation equation as defined inprevious sections of the study which will be
expounded depending on the case being studied.

Table 1 Percent differences of FFT frequencies inbeam cases
Sensor/Beam
Case
FL2

DATA PRESENTATION AND ANALYSIS

FL1

Sensors were based on the predetermined
location set to each sensor; FL2, FL1, L, M, R & FR
corresponds to Far Left 2, Far Left 1, Left, Middle,
Right and Far Right respectively. From theacceleration data, the time domain of theacceleration is transformed to a frequency domainusing the fast Fourier transform and the Welch
power spectral density analysis.
With regards to the Fast Fourier Frequency, theestimate acceleration response of the beam which
was taken from the Kalman filter code was subjectedto the process of determining the magnitude of theFast Fourier Frequency. Based from Fast Fourier
Frequency of all test trials, all frequency exhibit adominant frequency in the higher frequencies levels.
In each case, dominant frequencies of all trial fall ona specific frequency level. The sample data was thensubjected to an increase in the sampling rate to
reduce the amount of spectral leakage when a signal
is being filtered or processed. To further eliminate
the possible spectral leakage in the Fourier analysis,a process called Hanning windowing.

Fig. 8

L
M
R
FR

0”

1”

1.5
”

2”

2.5
”

3”

0.00
%
0.17
%
1.98
%
1.98
%
2.48
%
1.05
%

0.18
%
0.18
%
0.18
%
0.18
%
0.18
%
0.15
%

0.18
%
0.00
%
0.37
%
0.00
%
0.18
%
0.12
%

0.00
%
0.00
%
0.55
%
0.05
%
0.00
%
0.08
%

0.00
%
0.19
%
0.00
%
0.00
%
0.00
%
0.03
%

0.20
%
0.00
%
0.20
%
0.20
%
0.00
%
0.10
%

Using the same acceleration estimate obtained
from the impact hammer test which has been refinedby the Kalman filter, the PSD or the Power Spectral
Density of the beams response was established.
Similar to the FFT, a decline in the peak frequency
was also determined in the PSD and position of thesensors were also negligible.
Considering the damping ratio as a parameter toquantify and determine the damage severity of astructural element, the results of the analysis showthat the damping ratio of the beam decreases as theas the damaged found in the beam increases is anindication that the damage inflicted on the beam was
detected using the damping ratio as the parameter.
The severity of the damage was identified due to thefact that a continuous decrease in the damping ratio
was observed in all sensors. Although sensors
further away from the actual damage are less reliable
due to inconsistent decrease in damping ratio was
observed, sensors that are nearer the damage could
be noted to be more reliable and consistent.
Dissipation time however, increases as thedamaged sustained by the beam decreases. The
gradual increase in dissipation time go hand in handwith the gradual decrease in damping ratio due to theloss in the ability to remove the external force fromResampled FFT Frequency vs beam case

650

GEOMATE- Bangkok, Nov. 14-16, 2016

and the power spectral density remain constant in a
structural elements until its condition deteriorates
through damage. Besides the dissipation rate, theseparameters decrease depending on the damage
sustained by the said element. Larger decreaseindicate a severer damage whereas smaller changes
can be caused by smaller damages.as for thedissipation rate, damage is directly proportional tothe dissipation of energy in a structural element. By
evaluating any of these parameters, the damage wasquantified to determine its overall condition.

the beam. Results show that the sensor furthest fromthe actual damage tend to have more sporadic dataand is therefore less reliable as compared to theresult shown by the sensors near the cut.
In order to justify the completion of the study’s
objectives a summary of all results is tabulated toshow that the damaged was quantified and thatseveral methods can be used to determine thedamage severity levels of a beam or any otherstructural element being analyzed. Table 5.15
collected all the information from all previous
results of the study.

REFERENCES[1] Wang, S, Ren, Q & Qiao, P, “Structural
Damage Detection Using Local Damage
Factor”, Journal of Vibration and Control, Vol.
12, Sept. 2006, pp. 955-973.
[2] Yan, Y, Cheng, L, Wu, Z & Yam, L,“Development in Vibration-based Structural
Damage Detection Technique”, Mechanical
Systems and Signal Processing, Vol. 21, 2007,
pp. 2198–2211.
[3] Yoshida, I & Sato, T, “Health Monitoring
Algorithm by the Monte Carlo Filter Based on
Non-Gaussian Noise”, Journal of Natural
Disaster Science, Vol. 24, 2002, pp. 101-107.
[4] Sahin, M & Shenoi, R, “Quantification andLocalisation of Damage in Beam-like Structures
by Using Artificial Neural Networks with
Experimental
Validation”,
Engineering
Structures, Vol. 25, 2003, pp. 1785–1802.
[5] Ng, C, Veidt, M & Lam, H, “Probabilistic
Damage Characterisation in Beams Using
Guided Waves”, Procedia Engineering, Vol. 14,
2011, pp. 490–497.
[6] Zhu, X, Li, T, Zhao, Y & Yan, J, “Vibrational
Power Flow Analysis of Thin Cylindrical Shell
with a Circumferential Surface Crack”, Journalof Sound and Vibration, Vol. 302, 2007, pp.
332–349.
[7] Huh, Y, Chung, T, Lee, J & Kim, J, “Damage
Identification in Plates Using Vibratory Power
Estimated from Measured Accelerations”,
Journal of Sound and Vibration, Vol. 336, 2015,
pp. 106-131.
[8] Aldahdooh, M, Muhamad Bunnori, N & Megat
Johari, M, “Damage Evaluation of Reinforced
Concrete Beams with Varying Thickness Usingthe
Acoustic
Emission
Technique”,
Construction and Building Materials, Vol. 44,
2013, pp. 812–821.
[9] Nor, N, Ibrahim, A & Bunnori, N, “Diagnostic
of Fatigue Damage Severity on Reinforced
Concrete Beam Using Acoustic Emission
Technique”, Engineering Failure Analysis,
2014, pp. 1-9.

Table 2 Summary of Results
Beam Damage
Case
w/L
0.00%
0”
1.27%
1”
1.91%
1.5”
2.54%
2”
3.18%
2.5”
3.81%
3”
R-Value

FFT,
Hz

PSD,
Hz

Damping
Ratio

Rate

93.19

94.5850.28840.1451

87.47(6.14%)
85.52(8.23%)
84.93(8.86%)
81.22(12.84%)79.61(14.57%)87.318(7.68%)
85.144(9.98%)
84.778(10.37%)80.931(14.44%)79.237(16.23%)0.1915(33.60%)0.1911(33.74%)0.1531(46.91%)0.1110(61.51%)
0.0981(65.98%)0.1702(17.30%)0.1778(22.54%)
0.1982(36.60%)0.2442(68.30%)0.2769(90.83%)0.9804

0.97610.90230.9629

Dissipation

CONCLUSION
The study was able to establish a link between
the current condition of the beam to the uniquefrequency signature it emits using experimental
modal analysis such as the impact hammer test andthe use of filtering techniques such as the Kalmanfilter. Moreover, the damage severity of thestructural element, which in the study is a beam
element, was determined using the parameters suchas the dominant frequency of the frequency
signature, damping ratio and duration of damping.
Based from all the data gathered and analyzed fromthe study, the following observations and conclusion
were identified:
For determining the severity of the damage
sustained by the beam, sensor placement may be
negligible as seen in several analysis whereinposition of the sensor did not affect the results of thestudy. This could prove useful in monitoring thecondition of structural elements which are not easily
accessible for placement of measuring devices. This
however may not be the case for all parameters suchas the damping ratio wherein position was key tohaving an improved results.
Properties and parameter which include thedamping ratio, dissipation rate and dominantfrequencies found in both the fast Fourier transform651

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

STUDY DOWNTOWN STVANGER TO THE PEDESTRIAN SPACE
1

Tetsuya YANOBE1, Kazunari TANAKA2 and Shin YOSHIKAWA3Engineering, Osaka Institute of Technology, Japan; 2,3 Osaka Institute of Technology, Japan

ABSTRACTThe pedestrian space is substantial in downtown Stavanger. On the other hand, there are many big slopes inthe downtown of Stavanger. investigate the characteristic of the building and topography in the target space. The
authors make the characteristic of the city area using GIS. In this study, The authors clarified the influence of thepedestrian by watching the relationship between an element and the topography.
Keywords: Stavanger, Waterfront, Pedestrian, Slope
1. INTRODUCTION

2. CASE STUDY AREA STAVANGER

Norway is located in the Scandinavia peninsula
west coast in northern Europe. The shipbuilding
industry and the shipping trade prospered in Norway.
On the other hand, the shipbuilding industry
suddenly declined after the 1970s. Thus it becomes
the big problem in the city structure that waterfront
and a city area divide. After the 1980s, the city
reproduction project that used shipbuilding industry
ruins started. "A fjord city strategy" was devised in
2000. As a result, divided city area and waterfront
were unified. And the location of functions such as
shopping, eating and drinking, the residence and theculture was promoted in a shore part.
The pedestrian space is substantial in downtownof Stavanger. On the other hand, there are many big
slopes in the downtown of Stavanger. The authors
clarify the actual situation of the substantial
pedestrian space in a shore part and the city area.
The intention of this research was to clarify theactual situation of the pedestrian space paging
attention to the shape of building and topography.
Finally, a purpose of my study is to catch an
opportunity leading to the growth to find out leading
for regional by clarifying the actual situation of theboundary of spaces in the downtown Stavanger.

Stavanger is a Norwegian city and is the capital of
the Rogaland region. The area is about 71 km2. The
population is about 130,000 people. The history of
the town dates back to the Middle Ages, and old
buildings and new buildings exist together.
Stavanger is blessed with rich nature represented by
the fjord, and a pattern of the green tract of land anda waterside. The characteristic city is mixed.
In the downtown, a shopping area, a natural scene
district, bay area exist around a big pond.
Particularly, in the shopping area, the center does
well in composite facilities having a library and amovie theater. An open terrace characterizes many
restaurants. On the other hand, many place having
intense slant, which greatly affect the pedestrian.

Fig.１ Stavanger

Fig.２ Case study area
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GEOMATE- Bangkok, Nov. 14-16, 2016

3.A PURPOSE AND METHOD OF RESEARCH

5. THE CHARACTER OF BUILDING

In down town Stavanger, pedestrian space issubstantial. However, there are many big slopes inthe downtown of Stavanger. It is necessary to cancel
level differences between slope and store of the front
part of store in the slope, in front of various wall
surface forms and stores space exists. It is thoughtthat a lot of slopes influence pedestrian and a gettingout boundary of spaces in front of the store. I clarify
the actual situation of the boundary of space, which
paid its attention to the slope. As a summary, I
clarify the actual situation of the borderland by the
influence of the slope.
For the method of the study, the authors
investigate the shape of the building in the target
space and overflow of the store. The authors makethe characteristic of the city area using GIS. Because
there are many slopes in target area, The authors
calculate the road incline of the slope and clarify arelationship between the shape of building and theincline of the slope. As a result, the authors grasp
relationship between the characteristic of boundaryof space and incline of the slope.

The target area of this study is a downtown of theStavanger. A lot of shops exist in the downtown of
the Stavanger. In addition, the coast is full of tourists
because there are hotels, and many open cafés. This
space to my study is a doorway in the downtown.
The authors show a doorway to intend for at a red
point (Fig.4). The method was to collect data wasthree ways: observation, photo and note. The authors
photographed from No. 1 to No.150. And The
authors clarified from 151 to 389 by observation.
The authors finally summarized findings from No.1
to No.389 in a note. The authors show below amethod of listing (Tab.1).

4. BOUNDARY OF SPACES
The boundary of space is ambiguous space.
According to the previous study and the documents,
there are a Pilotis, an Arcade, an Atrium and a
Buffering space. My study looks for an existingboundary of space between public space and private
space. I define the public space in the site or theprivate space outside the site as a boundary of space.

Fig.4 Target area

No.
１
２
３

Tab.1 The method of listing
The type in front
The typeelementsof the shopof building
IS typeApparelIS type
OthersBillboard
Pocket type
ApparelBillboard

6. THE ROAD INCLINE OF THE SLOPE
For the method of the road incline of the slope,Firstly, The authors understand the spot where
height changes by Google Earth. And, the authors
find the distance of each change point by Google
Earth and calculate the road incline of the slope. The
authors show a calculating formula in Fig.5.

Japan𝜃𝜃 = 𝑡𝑡𝑡𝑡𝑡𝑡−1

NorwayFig.5

Fig.3 In lecture to extension of boundary space653(𝑦𝑦2 − 𝑦𝑦1 )(𝑥𝑥2 − 𝑥𝑥1 )The calculate of incline

GEOMATE- Bangkok, Nov. 14-16, 2016

7. THE TYPE OF BUILDING AND ELEMENTS

8. RELATIONSHIP BETWEEN A BUILDINGSHAPE AND THE ROAD INCLINE

The authors investigated the shape of the building,the type of the store and the kind of the element forfrom September 2015 to October 2015. And theauthors paid my attention to the road incline.
The authors show the shape of building, the typeof the shop and the kind of the element in the map(Fig.6.7.8). The authors made clear that a pocket
type, stairs type, a restaurant, an apparel shop and ageneral store occupied most in the building of thedowntown. And the authors clarified many products
are exhibited in the south side of the downtown. The
authors understood the change of the building shapeevery street. It is thought that eaves type and thestairs-shaped street come under other influences.
The authors paid my attention to the road incline.

The authors hypothesized that the shape of thebuilding might change by the slant of the slope. The
authors calculated the road incline of the slope byeach building unit and grasped the relationship withthe building shape. The target street is 6 streets
which roads incline has a big and a small street in
total.
As a result, there is much roof type of buildings at
the road incline less than 4% more than 0%. On theother hand, the authors clarified that stairs type ofbuildings was strong at the incline less than 8%
more than 4%. Stairs type is more likely to beinstalled in the doorway as the shape that apedestrian is easy to go in and out of the building
when a road incline is big. On the other hand, it isthought that there are many eaves types that set up
open cafe and planting when a road incline is small.
As a result of having paid its attention to a road
incline, I clarified the relationship between the shapeof the building doorway and the outdoor spaceelement.

Fig.6 The shape of building
Fig.9 The target street
Composition ratio (%)

Fig.7 The type of shop
It is less than 1% more than 0%

Road incline (%)

Fig.10 The result of analysis

Fig.8 The kind of element654

GEOMATE- Bangkok, Nov. 14-16, 2016

9.ANALYSIS AND DISCUSSION OF RESULTS
I clarified that the type that accounted for most of
the target area was roof type and a pocket type andstairs type. Moreover, I made clear that a billboard,planting, an open cafe, a goods accounted for mostas for the existing element in a boundary of space. I
clarified relationship between the element in theborderland and the shape in front of the building. A
lot of planting and open cafes existed in the roofshaped building. The variety of elements existed in
the pocket-shaped building. Thus, I thought that thepocket-shaped building had high borderland degree.
By these investigations and analyses, I was able toget knowledge about a connection of the public
space and private space in Stavanger. Depending ona shape and the slant in front of the building, it isnecessary for the people of the store to place theelements such as products in front of a shop to use
this knowledge in practice. I strongly suggest it tothe owner of the building.

outside open space typeslope typeroof typestairs type

10. CONCLUSION
outside open space typeAs a result of having arranged the type of the shop
and the data of the shape of the building, The
authors clarified the kind of the element and therelations of the position. The authors understood theinfluence of the pedestrian by watching therelationship between an element and the topography.
The authors suggest the space that is easy to walk insidewalk for a pedestrian by considering therelationship between the characteristic of the
sidewalk and a shape of the building.
I think when the boundary of space changes bytime of the day. Thus, I will find new technique tomake a boundary of space in future by paying its
attention to a change at time, and analyzing aboundary of space. In future, the authors will
suggest the space that is easy to come and go to thebuilding for a walker by considering the relationshipof the characteristic of the sidewalk and a shape ofthe building in the downtown of Stavanger.

・The method of pedestrian behavior listing
A purpose of the investigations is to clarify
whether a pedestrian pays his/her attention to whatkind of building, element. This table is a listing
example of the observation result.

REFERENCES[1] Hori Masahiro : Sustainable urban developmentin North Europe three countries Private urban
development promotion mechanism
Urban Study Vol.55 2012
[2] Ashihara Yoshinobu: Aesthetics of thecityscape[3] Space study subcommittee : Space education
celebration for a building, city planning
Inoueshoin

ANNEX

passage typenone typepocket type655

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

ON THE RELATIONSHIP BETWEEN THE SPATIAL ELEMENTS AND
PERCEPTUAL SPACE OF CHILDREN
1

Takumi SAKAI1, Kazunari TANAKA2 and Shin YOSHIKAWA3Engineering, Osaka Institute of Technology, Japan; 2,3 Osaka Institute of Technology, Japan

ABSTRACT: It is realized the Comprehensive Support System for Children and Child-rearing at many cities inJapan. On the other hand, this idea is that guardian is a subject that the child cares. It is important that therelationship of the environment for the child and city planning. The authors understand the cognitive space ofchildren by the questionnaire survey. And the authors analyzed the relationship between the shape of town andreality space by Geographic Information System. The authors measure the distance between objects by network
distance.
Keywords: Cognitive Map, Cognitive Space, Distortion
1.

INTRODUCTION

2. A PURPOSE AND METHOD OFRESEARCH

Our country makes the towns that established an
important point for the childcare support as
declining birthrate measures. However, there are few
places where children play in city space. The current
city planning does not become the child-based
measures. Thus, it is necessary to plan a city after
having grasped the relationship between a children
and the town.
In this study, the authors analyze a relationshipbetween the cognitive space of the children andreality space. In previous of study, the authors
analyzed the element that gave a distortion on a map
by grasping a difference of the cognitive space andreality space as a distortion. As a result, the authors
understood the distortion of the cognitive space isdistance and an angle [1]. The authors pay attention
to a school road and a road incline as a kind of
distortion elements. The authors analyze image
amplification and the spatial perception of the target
space.

A purpose of this research is to grasp space and afactor to feel that a child is attractive by analyzing
relationship between the cognitive space of thechildren and reality space. The authors grasp a
distortion of the cognitive space from a psychology
side and physical aspect. To do that, it is necessary
to externalize cognitive space. There are threemethod of externalization (1:Map Drawing Method,
2:Multi Dimensional Scaling, 3:Cognitive Distance
Estimation method) [2]-[4]. It is necessary toconsider the characteristic of research participants
and research conditions. The map drawing method issuitable for the case for children than the cognitive
distance estimate method to require abstract work[5]. In this study, the authors acquire the cognitive
map by conducting a questionnaire survey (1:Map
Drawing Method). In the research method, thequestionaries’ survey to confirm the cognitive spaceof the children is performed. The authors calculate
the amplification of the space image as a standard of
the distance in a school road. The authors measure
the distance using network distance. It is consideredthat the information of the height direction andanalyze the cognitive space.
3.

Fig.1

THE TARGET AREA

The target area of this research is an elementary
school and outskirts area in Osaka (Fig.2). The
precinct range is north-south 800m, east-west
1,600m. There are many slopes in the precinct range.
The pedestrian moves to the height direction in
addition to a plane. So, it is important that theauthors understand the distortion of the height
direction.

Amusement place of children

656

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig.4

Fig.2 Precinct range
4.

The method of calculation

THE QUESTIONAIES’S SURVEY

The authors investigated cognitive map survey
from September 2014 to October 2014. The subject
is a 5th grader and a 6th grader (83students) [6],[7].
The authors order subjects that can write the town
where you lived in (Fig.3). When a subject
surpassed a range of the distribution paper, theauthors performed addition of the paper.

Fig.5
6.
Fig.3
5.

The calculation results

THE POSITION OF SPACE

The authors measure the distance between eachspace and home or schools to clarify the relationshipbetween an image amplification of the space and theposition. The distance measurement is not distancein a straight line between two points (Euclidean
distance). The authors use network distance. So, theauthors can perform analysis in consideration of theeveryday life of the children. In this study, theauthors classify space as the outside of the schoolroad and inside of the school road. And the authorsclassify the home side (A) and the school side (G) in
outside the school road. On the other hands, theauthors classify in front of the home side (B), in
front of the school side (F), home side is based on
network distance (C), center area (D), school side isbased on network distance (E) in inside the school
road.
Because the cognitive map is described into ahalftone plate form, it is necessary to make astandard to judge the inside and outside. If distanceof home and a school and the each space did notexceed the shortest course distance of the school
road, the authors assumed it with the inside school(Fig.6).

Cognitive map

THE IMAGE AMPLIFICATIONThe range to use well of the cognitive map of thechildren is drawn (school road, home etc.). Many
elements are drawn on the imminent space.
Therefore, The authors calculate amplification of
the space by comparing the distance of the road withthe reality space and the cognitive map. Not only the
distance of the road but also the element in the spaceis considered. The amplification that the authors
calculated assumes it the amplification of the imageof the space. There are not a scale and a direction ona cognitive map. Therefore, it is necessary to arrange
the standard line to compare it with the reality space
on a cognitive map. In this study, the authors assume
the school road as a kind of the standard line (Fig.4).
As a result, the authors grasped that the spacearound the home and the school extend (Fig.5). So,
the authors clarified that many children can write theplace that usually lives well. On the other hands, thesubject did not write the road that passed without
thinking about anything.

657

GEOMATE- Bangkok, Nov. 14-16, 2016

Tab.1
A
B
C
D
E
F
G
Fig.6

Fig.7

MedianMore than 1Less than 1

0.964
1.341
1.160
1.128
1.321
1.473
0.769

0.816
1.292
1.031
0.961
1.130
0.936
0.753

14
13
40
52
24
13
5

26
7
37
62
15
15
33

As a result, the space that spread is approximately
2 times than the space that reduced. Because themedian doubles 1.29, the authors can grasp whatchildren greatly image in cognitive space. However,there is the median becomes the value that is almost
1.000. So, the authors clarified existence of the child
who greatly had the image of the school and thechild who had small (P-value < 0.01).
The medians of the home side and the school side
are less than 1.000 in outside of school road. The
authors understand a cognitive map image tends tobe reduced. Therefore, Image distance becomes
small in the outside of the school road. The authors
can grasp a thing that is not important.
The median of the home side and the central space
are the values that are almost 1.000. And there are
not the differences between reduction space andextended space. However, there are differences
between reduction space and extended space in theschool side. Thus, the authors clarified existence ofelements that children is easy to recognize.

The judgment of inside and outside

The calculation e.g. of the space distance

7. THE RELATIONSHIP BETWEEN IMAGEAMPLIFICATION AND POSITION
The authors grasped relationship between aposition of each space and the amplification of theimage (Fig.8). The home position and an elementary
school are written in acquired 83 maps. Therefore,
the author used the samples by the 20 maps thatthese lead to on a road. And the authors indicated theaverage and the median, the number of the spacethat spread and the number of the space that reduced(Tab.1). A difference occurred to the mean and the
median. Because the cognitive map has the map that
subject extremely describe. So, the authors use themedian to clarify the characteristic of each space.

Fig.8

Classification result

Average

8.

THE DISTORTION OF THE HEIGHTThe children wrote the height information such as
the slope on a cognitive map. The authors thought
this information affects the distortion of thecognitive space. Thus, the authors investigatedcognitive map survey about the distortion of theheight on March 2016 (Fig.9). The subject is 10
people in 50s from the teens (1241points data). As aresult, the authors classified the relationship between
height of road incline and distortion of the height on
cognitive space (Fig.10).

The result of analysis
Fig.9
658

The questionnaireGEOMATE- Bangkok, Nov. 14-16, 2016

Fig.11
Fig.10The cognitive difference (distance, incline)

10.

The children play on the road

REFERENCES

9. CONCLUSION[1] Sakai T, Tanaka K, Yoshikawa S “The study of
distortion on children’s cognitive map”JSCEKC Ⅳ-４ 2015
[2] Lynch K, “The Image of the City”, MIT Press[3] Masui Y, Imada H, “Methods andmethodological issues in the externalization of
the cognitive maps”, 42(2), pp.65-81,
humanities review 1992[4] Okamoto K, “A Review of Cognitive DistanceStudies”, Vol.34 No.5, pp.429-448, Japanese
Journal of human geography 1982
[5] Wakabayashi Y, “The spatial analysis ofcognitive map”, pp.1-318, Tijin Shobo, 1999
[6] Murata K “ The map of children’ world-Secretbase・Children’s route・The space that a
haunted house forms”, pp.1-176, ReimeiShobo,
1988[7] J. Piaget “A condition and learning of thedevelopment”, pp.1-219, Seishin Shobo, 1979
[8] Kaneda H, Nakagawa T, Biwa Y, Tanaka K,
Yoshikawa S, “The pedestrian space that paid
its attention to the viewpoint of the children”,
JSCE-KC Ⅳ-47 2013
[9] Umekawa K, Yokoyama H, Miyagishi Y “A
study on the playgrounds for the children andthe space cognition in the city”, AIJ-Kinki
Branch, pp.365-368, 2009

In this study, the authors caught a distortion of thespace as image of the children. The authors clarified
the space that is important to children by calculating
image amplification. And the authors understood thecharacteristic of children’s cognitive space byclarifying the relationship between the amplificationand position.
In addition, the authors classified that the
distortion of the height affected a road incline anddistance by investigating the distortion of the height.
We, urban designers plan the parks using theformal form. However, children plays everywhere
we do not image normaly, like a part of road, a river,
big trees in cities and so on (Fig.11). They are notprepare by urban designers. There were also thespecial places for children. [8] In the study of
Umekawa [9], the children do not choose the place
to play by its size.
As a result of this study, it is clear that thecognition sizes of the spaces by children is different
where they are. We, the designers have thepossibilities to design new place for children in real
space.
In the future, the authors will analyze therelationship between the spaces that shows frequent
use and the position of the drawing element.

659

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

CHARACTERISTICS OF COLLAPSE TIME OF EARTHQUAKESOURCE ROCK AS ONE PARAMETER TO PREDICTION THE
EARTHQUAKE IN YOGYAKARTA REGION
Hita Pandita1), Sukartono2) and Agustinus Isjudarto3)
Department of Geology, Sekolah Tinggi Teknologi Nasional, Yogyakarta, e-mail: hita@indo.net.id
Department of Geology, Sekolah Tinggi Teknologi Nasional, Yogyakarta, email: kartono_skt@yahoo.co.id
Department of Mining Engineering, Sekolah Tinggi Teknologi Nasional, Yogyakarta, email: is_darto@yahoo.com
AbstractIndonesia is a country with high potential to get an earthquake, one of them is Yogyakarta and surrounding
areas. Although earthquake attack in Yogyakarta area have been several times, efforts to reduce the number ofvictims is dificult to do. It’s cause the absence of parameters that can be used to estimate the arrival time of theearthquake. One of the parameters that may be used is predicted when a fault will move on, using an earthquake
if get an increase of pressure constantly. This study aimed to find out when a rock will collapse if it get a variedpressures, so it can get a collapse rock time formula. If the formula is known, it can be modeled to the load
received a fracture for understand when the fracture will move on. The research method used is a field
investigation, and rocks compressive strength test. By using the test six variations of loading finally made it
known pattern of the outbreak of the rock. Formulation successfully arranged by using four models of the curve,
namely: linear, logarithmic, exponential and quadratic functions (power). Three types of rocks have been tested,
and the results show the most likely around earthquake source rocks around Opak fault is a diorite rock, since
the collapse of this rock closest to the timeframe of seismicity induced by Opak fault.
Keywords: Prediction, Earthquake, Strength test, Time, Opak fault

INTRODUCTION
Yogyakarta is an area with high potential
earthquake.
A number of large magnitude
earthquake hit the region like in 1867 and 2006 [1].
Damage and casualties from those earthquake isvery large. The large number of souls who drift andloss of property caused by the lack of accurate
predictions about when an earthquake will occur.
According from many geologist, most of theearthquake is cause by sudden release of strainenergy that has built up over a period of time [2].
The pressure on the earth's crust in the earthquake
areas are generally caused by the collision between
two or more plates or continents and oceans. Basedon the theory of occurrence of the earthquake, thenthe expected arrival of an earthquake in a region
can be estimated from the time exceeding of the
durability ofrocks in the earth's crust. So
understanding the durability of the rocks in theearth's crust will be able to help improve theaccuracy of prediction of the earthquake in theregion.
This study aimed to find out when a rock will
collapse if it get a varied pressures, so it can get acollapse rock time formula. If the formula isunderstand, it can be modeled to the load received afracture for understand when the fracture will moveon. The research object is rocks which predicted as
earthquake source around Opak Fault in
Yogyakarta area. The choice of Opak Fault
becaused many great earthquake occurred aroundthis fault [1].
METHODThe method used in this research is the fieldsurvey, stratigraphic analysis and compressive
strength test. Field survey carried out in theSouthern Mountain to determine the thickness of
sedimentary rocks and understanding of earthquake
source rock. The field investigation also taken rocksamples were estimated as earthquake source rocks.
Laboratory investigations carried out in the form of
uniaxial compressive strength test of theearthquake source rocks.
REGIONAL GEOLOGY
The research areas included in border between
the Southern Mountains of East Java Zone andYogyakarta Depression (van Bemmelen, 1949) [3].
The boundary between these two physiographic
zones by some researchers suspected a Northeast660

GEOMATE- Bangkok, Nov. 14-16, 2016

Southwest Fault known as the Opak Fault
Depression Yogyakarta is a lowland filled bydeposition of volcanic activity of Merapi. Southern
Mountains of East Java is a high land stretchesfrom the eastern part of Yogyakarta to thesoutheastern tip of the island of Java.
Regional stratigraphy research area has beenreviewed by several researchers such as Rahardjo,
et al (1995) [4], Surono, et al. (1992) [5] and van
Bemmelen (1949) [3]. The thickness of
sedimentary rocks in the Southern Alps isestimated at more than 5,000 m (Pandita &Sukartono, 2014) [6], the bedrock is estimated to(Rahardjo,
et
al,
1995)
[4].
be
Cretaceous
metamorphic
complex.
Unconformity at top of the bedrock deposited
limestone from Wungkal Formation at Eocene.
Series of volcanic deposits began to appear at
Kebo-Butak Formation begins at Olgosen (Surono,2008) [7]. Then sequentially formed Semilir
Formation and Nglanggran Formation the Early
Miocene to Middle Miocene. In the early Late
Miocene volcanic series is gradualy being replaced
by a series of carbonate rocks, beginning by
Sambipitu Formation and ending by Kepek
Formation (Pandita, et al, 2014) [6].

Fig1. Samples location BYT=Bayat, dan JB = Jember

RESULT
Material And Location
Based on the study of the researchers
mentioned above, then there are three possible
earthquake source rocks around Opak fault, namely
the continental crust rocks (granite), metamorphic
complex and diorite. According to these, there were
three locations selected for sampling. Twolocations of the Bayat region, namely in the Pendul
hill (BYT14) and Jokotuo (BYT10), whereas one
location area Rajegwesi, Jember (Figure 1). At
BYT10 of samples taken in the form of marble
rocks of the complex metamorf Bayat. At BYT14
samples taken form of diorite rock. Location
Rajegwesi an analog samples of the continental
crust in the form of dacite rock.

Sampling locations for compressive strength
test selected based on the type of rock to be tested.
Sampling in the form of surface samples fromsome rocks that are suspected as earthquake source
rocks. Pandita, et al. (2014) [6] and Prasetyadi(2007) [8] based on stratigraphic studies predictthat the bedrock of the Southern Mountains aremetamorphic complex. Santosa (2014) [9] is basedon the gravity study estimated bedrock Southern
Mountains is characterized like diorite rocks.
Sribudiyani, et al. (2003) [10] and Smyth, et al(2007) [11] estimates that the bedrock of East Javais the continental crust.

661

GEOMATE- Bangkok, Nov. 14-16, 2016

Because x represents the magnitude of loading,whereas the compressive strength test loading has
notation v, then a notation "x" changed to "v".
Likewise with "y" which is the time, and in a general
sense the notation is "t", so the notation "y" replacedwith "". Thus the formula can be changed to:
Marble:t = 793.3v-1.65 4)
Dacite:t = 1266.v-1.90 5)
Diorite:t = 342.9.v-1.02 6)

Uniaxial Test
Three rocks prepared for uniaxial test, witheach rock type was prepared in six different loading.
Samples were prepared in the size of 5x5x5 cm
appropriate standard test ISO 2825: 2008 [12]. Tests
carried out at the Laboratory of Materials andConstruction Engineering at the Faculty of Civil
Engineering UII Yogyakarta. Condition of the sample
in a dry state. Compressive strength testing is donewith six different loading speeds, namely: 1 kN/s, 2
kN/s, 3 kN/s, 4 kN/s, 5 kN/s and 6 kN/s. Full resultsof the compressive strength test can be seen in Table
1.

Table 1. Result of uniaxial test from three rocks
Analysis
Data processing was performed using Excel
2007 program to get formula using Power curve(quadratic function). Power curve (quadratic
function) is the best results were obtained through aseries of processing from the other curves (Figure 2).
The curve shows the X-axis as speed of loading in
kN/sec, the Y axis is the collapse time of the rocks ina seconds.
Based on the power curve (quadratic function)
obtained formula the relationship between speed ofloading with the collapse time of the rock. The
formula are (Fig. 2):
1)
Marble:y = 793.3x-1.65
2)Dacite:y = 1266.x-1.90
3)
Diorite:y = 342.9.x-1.02

Figure 2. Quadratic function curve relationshipbetween variations in the speed of loadwith the collapse times of three rocks
results from processing using Excel
program.
DISCUSSION
Force of Collision
Until now there has been no study that
calculates the magnitude of the force or load that
occurs as a result of plate collision in the southern
island of Java. To get these formulations then were
calculated theoretically based on the generalcharacter of the type and speed of plate. Calculation
662

GEOMATE- Bangkok, Nov. 14-16, 2016of two types, namely the imposition of macro andmicro load. The macro loading according to its
natural condition based on area of collision in theplane of Opak fault. Imposition of a micro issimulation if the size of the collision area
corresponding scale of laboratory test. For the
calculation required data of the volume of plate,
speed of plate movement and density.
On a macro scale counting the length of theplate is calculated from the collision point to the
spreading point, based on a calculation from theGoogle Earth obtained 2790.97 km. Width has
calculated from the collision area with the Opak fault,
ie 21.35 km. The thickness is calculated based on the
average thickness of the oceanic plates, which is 5
km. The final result of the volume can be seen in
Table 2..

Earthquake Prediction
Formulas of the collpase time of each rock will
use to calculate the earthquake prediction.
Calculations performed by a micro scale load, this isdue to micro-scale approach the area on a scalelaboratory tests. Value of v is taken from the amountof pressure that occurs in kN/sec, while t is the timeof earthquake prediction searched in seconds.
The result of calculation shows that dacite rocktakes more than 7,800 years to get an earthquake,marble rocks need 400 years and diorite rocks need
173 years (Table 3). The study on the history ofseismicity around Opak fault, there are two major
earthquakes that in 1867 and 2006. This represents aninterval of 139 years between the first quake toward
the second quake. Based on that show diorite rocks
closest to the earthquake time interval. So there is agreat possibility the earthquake source rock is diorite
rocks.

Table 3. Calculation time prediction of an earthquake
on Opak fault of the three types of seismic
source rocks.
Table 2. The results of the theoretical
calculation of geometry and its burden on the IndoAustralian plate movement against geometry of Opakfault

Seeing the time difference between simulation
on rock diorite with intervals of earthquake that
occurred on Opak fault, it is likely due to several
things:1) Opak Fault is already a weak field, so it does
not require loading large when compared withthe condition before the fracture.
2) The possibility of an earthquake source rocks
have lower resistance values of diorite.
3) There are many assumptions in this study likethe volume of the Indo-Australian plate, thedensity of which is taken from the averagevalue oceanic plate, thus affecting the value ofthe load.

Calculation of macro-scale load can becalculated by knowing the average speed of the IndoAustralian by 6 cm / yr (Whittaker et al., 2007) [13]or 1.9 x10-7 cm/sec. By using an average density
oceanic plates 3.3 gr/cm3. It can be seen from theplate macro load of 1870.9 kN/sec.
Because in this study did not test the sample size
variable, then the load on a macro scale can not be
used to calculate the time of rupture of rocks. In thisconnection it is necessary to micro-scale modeling or
calculation in accordance with a surface area in thelaboratory tests. On a micro scale modeling of somevalue on a macro scale still, the only difference being
the volume due to changes in the loading area. The
final result obtained at the micro scale imposition of-5
4.38 x 10 kN/sec (Table 2).

CONCLUSION
This research is still an early stage in an effort
to prediction of earthquake in the future in the
Yogyakarta area. However, this study has shown thepossibility of the arrival of an earthquake can be663

GEOMATE- Bangkok, Nov. 14-16, 2016

expected from collapse time of the earthquake source
rocks. This research still needs to be improved, with avariety of other rock samples and also variations inloading time, so that the preparation of the formula
becomes more accurate.

[6] Pandita, H. & Sukartono, “Karakteristik Nilai
Kuat Tekan Batuan Sumber Gempa Sebagai
Salah Satu Parameter Penentuan Terjadinya
Gempa Bumi di Wilayah Yogyakarta”, Research
Report, STTNAS, 2014, pp. 53.
[7] Surono, “Litostratigrafi dan Sedimentasi Formasi
Kebo dan Formasi Butak di Pegunungan
Baturagung, Jawa Tengah Bagian Selatan”, Jurnal
Geologi Indonesia, Vol. 3 No. 4 Desember 2008:
183-193[8] Prasetyadi, C., “Evolusi Tektonik Paleogen Jawa
Bagian Timur”, Doctoral Thesis, Unpublished,
ITB, 2007.
[9] Santosa, A., “Determination of Ancient Volcanic
Eruption Centre Based on Gravity Methods in
Gunungkidul
Area
Yogyakarta”,
Proc.
International Symposium of Geoheritage, UPN,
2014. pp[10] Sribudiyani, Muchsin,N., Ryacudu, R., Kunto,
T., Astono, P., Prasetya, I., Sapiie, B., Asikin,
S., Harsolumakso, A.H., & Yulianto, I., “The
Collision of the East Java Microplate and Its
Implication for Hydrocarbon Occurences in EastJava Basin”, Proc. Ann. Conv. IPA 29th, Jakarta,
2003, pp 1-12
[11] Smyth, H.R., Hamilton, P.J., Hall, R., & Kinny,
P.D., “The deep crust beneath island arcs:
Inherited zircons reveal a Gondwana continental
fragment beneath East Java, Indonesia”, Earthand Planetary Science Letters, Science Direct,
Elsevier, 2007, pp 269-282.
[12] SNI, Cara Uji Kuat Tekan Batu Uniaksial,
Badan Standardisasi Nasional, 2008.
[13] Whittaker, J.M., Muller, R.D., Sdrolias, M., &Heine, C., “Sunda-Java trench kinematics, slab
window formation and overriding platedeformation since the Cretaceous”, Earth andPlanetary Science Letter, 255, Science Direct,
Elsevier, 2007, pp 445-457.

ACKNOWLEDGEMENT
This study was established with funds from the
Ministry of Research and Higher Education, in theform of competitive grants during 2014 and 2015,
with the contract number SPDIPA-023.04.2.189971 /2014 to 2014 and SPDIPA-023.04.1.673453 / 2015
for funding in 2015. Thanks are expressed to thepaleontology laboratory assistants who assist in datacollection in the field.
REFERENCES[1] Thant, M., Kawase, H., Pramumijoyo, S.,
Hendrayana, H., & Adi, A. D., “Focal Mechnismsof Subduction Zone Earthquakes Along The Java
Trench: Preliminary Study For The PSHA For
Yogyakarta Region, Indonesia”, The 14th World
Conference on Earthquake Engineering October
12-17, Beijing, China. 2008,[2] Kearay, P., Klepeis, K.A., & Vine, F., Global
Tectonics, 3rd ed., Wiley-Blackwell, A John
Wiley & Sons, Ltd, 2009, pp 92-94.
[3] Van Bemmelen, The Geology of Indonesia. The
Hague, Martinus Nijhoff, vol. IA. 1949, p. 546658[4] Rahardjo, W., Sukandarrumidi, & Rosidi, H.M.D.,
Geological Map of Yogyakarta Sheet, Jawa,
Geological Research And Development Centre,
Bandung, 1995.
[5] Surono, Toha B., Sudarno I., & Wiryosujono, S.,
Geological Map of Surakarta and Giritontro
Jawa, GRDC Bandung, 1992.

664

Sixth International Conference on Geotechnique, Construction Materials andEnvironment,
Nov.2016
14-16, 2016, ISBN: 978-4-9905958-6-9 C3051
GEOMATE- Bangkok,
Bangkok,Thailand,
Nov. 14-16,

MATHEMATICAL MODELING OF FOREST FIRES INITIATION, SPREAD
AND IMPACT ON ENVIRONMENT
Valeriy Perminov1, Alexander Goudov2
Department of Ecology and Basic Safety, Tomsk Polytechnic University, Russia
2
Mathematical Faculty, Kemerovo State University, Russia

1

ABSTRACT
A mathematical model of surface and crown forest fires spread and impact is considered. A three dimensional multiphase,
physic based model is used. The boundary-value problem is solved numerically by finite volume method. This model has
been applied to describe the process of initiation and spread of surfaces forest fires and their transfer into crown of forest fires.
The results of numerical solutions present the distribution of the main functions of the process (the velocity field, thetemperature of gas and solid phase, the concentration of the oxygen, gas product of pyrolysis and inert components, etc.) overtime. Scenarios modeled within this study represent a possible approach to the preliminary assessment of risk and should beverified by more detailed CFD modeling.
Keywords: Crown Fire, Surface Fire, Mathematical Modeling, Finite Volume Method
INTRODUCTION

heterogeneous medium; 2) in the fire zone the forest is aporous-dispersed, two-temperature, single-velocity, reactive
medium; 3) the forest canopy is supposed to be nondeformed medium (trunks, large branches, small twigs andneedles), affects only the magnitude of the force ofresistance in the equation of conservation of momentum inthe gas phase, i.e., the medium is assumed to be quasi-solid(almost non-deformable during wind gusts); 4) let there be a
so-called “ventilated” forest massif, in which the volume offractions of condensed forest fuel phases, consisting of dry
organic matter, water in liquid state, solid pyrolysis products,
and ash, can be neglected compared to the volume fractionof gas phase (components of air and gaseous pyrolysis
products); 5) the flow has a developed turbulent nature andmolecular transfer is neglected; 6) gaseous phase density
doesn’t depend on the pressure because of the low velocitiesof the flow in comparison with the velocity of the sound.
Let the coordinate reference point x1, x2 , x3= 0 be situated at
the center of the domain of surface forest fire source at theheight of the roughness level, axis 0x1 directed parallel tothe Earth’s surface to the right in the direction of theunperturbed wind speed, axis 0x2 directed perpendicular to
0x1 and axis 0x3 directed upward (Figure 1).

One of the objectives of these studies is the
improvement of knowledge on the fundamental physicalmechanisms that control forest fires initiation and spread. A
great deal of work has been done on the theoretical problem
of how forest fire initiation. In forest there are two steps forcrown forest fire initiation: spread of fire from crow tocrown and crown fires are initiated by convective andradiative heat transfer from surface fires. However,convection is the main heat transfer mechanism. Firstly,
crown forest fire initiation has been studied and modeled by
Van Wagner [1]. There are three simple crown properties:
crown base height, bulk density and moisture content offorest fuel in this theory. Also crown fire initiation havebeen studied and modeled in detail (eg: Alexander [2], Van
Wagner [3], Xanthopoulos, [4], Rothermel [5,6], Van
Wagner, [7], Cruz [8], Albini [9], Scott, J. H. and Reinhardt,
E. D. [10]. The discussion of the problem of modeling forest
fires is provided by a group of co-workers at Tomsk
University (Grishin [11], Grishin and Perminov [12]. The
general mathematical model of forest fires was obtained by
Grishin [11] based on an analysis of known and original
experimental data [11,13], and using concepts and methods
from reactive media mechanics. The physical two-phasemodels used in [14-15] may be considered as a continuationand extension of the formulation proposed by Grishin andPerminov [12]. However, the investigation of crown fires
has been limited mainly to cases studied of forest fires
initiation in two dimensional settings and did not take into
account space properties of these phenomena.
PHYSICAL AND MATHEMATICAL MODEL
It is assumed that the forest during a forest fire can bemodeled as 1) a multi-phase, multistoried, spatiallyFig. 1.

665

Schematic of a forest fire domain.

GEOMATE- Bangkok, Nov. 14-16, 2016x2  x2 e :Using the results of [11-12] and known experimental
data [13] we have the following sufficiently generalequations, which define the state of the medium in the forest
fire zone, written using tensor notation.

(  v j ) m , j  1,2,3, i  1,2,3;
t  xjdvP i (  vivj )  scd vi | v |  xi  x jdt
 vi ; g i  mdT( c p vj T )  q5 R5   v (T  Ts ) c p
dt  x j

x3  0 : v1  0, v2  0,(1)

  c U R 
 kcU R  4k S TS4 x j  3k x j 
 4k g T 4  0, k  k g  k S ,4T i c pi  i S  k s (cU R  4TS4 ) 

t
i 1
 q3 R3  q 2 R2   V (T  TS );

v
c
 v1
v
 0, 2  0, 3  0,   0,
 x3
 x3
 x3
 x3
T
c  UR c 0, UR  0 .
3k  x3 2
 x3

x3  x3e :
(3)

(4)

(5)

3

(7)

c  1, p  RT   , v  (v ,v , v ).

M


c

e

1



1 2

3

1

The system of equations (1)–(7) must be solved taking
into account the initial and boundary conditions:t  0 : v1  0, v 2  0, v3  0, T  Te ,c  ce ,Ts  Tse ,  i   ie ;x1   x1e :v1  Ve , v2  0, v3  0, T  Te , c  ce ,c U R с U R  0,
3k x1 2
v
 c
v
 v2
x1  x1e : 1  0,
 0, 3  0,
 0,
 x1
 x1
 x1
 x1
T
c  UR c 0, UR  0 ; x1
3k  x1 2

(8)
(9)



 v3
c
 v1
 v2
 0,
 0,
 0,   0,
 x2
 x2
 x2
 x2
T
c  UR c 0, 
 UR  0 ; x2
3k  x2 2

x2  x20 :

(13)

(14)

d
is the symbol of the total (substantial)
dt
derivative: d    v   v   v  , v is the123dt  t
 x1
 x2
 x3
coefficient of phase exchange;  - density of gas – dispersed
phase, t is time; vi - the velocity components; T, TS, temperatures of gas and solid phases, UR - density ofradiation energy, k - coefficient of radiation attenuation, P pressure; cp – constant pressure specific heat of the gas
phase, cpi, i, i – specific heat, density and volume offraction of condensed phase (1 – dry organic substance, 2 –moisture, 3 – condensed pyrolysis products, 4 – mineral partof forest fuel), Ri – the mass rates of chemical reactions, qi –thermal effects of chemical reactions; kg , kS - radiation
absorption coefficients for gas and condensed phases; Te the ambient temperature; c - mass concentrations of  component of gas - dispersed medium, index =1,2,3,
where 1 corresponds to the density of oxygen, 2 - to carbon
monoxide CO, 3 - to carbon dioxide and inert componentsof air; R – universal gas constant; M , MC, and M molecularmass of  -components of the gas phase, carbon and air
mixture; g is the gravity acceleration; cd is an empiricalcoefficient of the resistance of the vegetation, s is the
specific surface of the forest fuel in the given forest stratum.
To define source terms which characterize inflow (outflow
of mass) in a volume unit of the gas-dispersed phase, thefollowing formulae were used for the rate of formulation of , outflow of oxygen R51 andthe gas-dispersed mixture m

Here and above

1

3

(12)

v3  0, T  Te , x1  , x2   ;

(2)

(6)

 1
 2
  R1 ,  2
  R2 ,
t
t
M

 4
 0;
 3 3   C R1  C R3 ,  4
M1
t
t

 c
c  UR c 0, 
 UR  0 , x3
3k  x3 2

v3  v30 , T  Tg , x1  , x2  

 k g (cU R  4T 4 )dc c ,   1,3;
  (  v j c )  R5  m
dt
 xj

v
c
 v1
v
 0, 2  0, 3  0,   0,
 x2
 x2
 x2
 x2
T
c  UR c 0, UR  0 .
 x2
3k  x2 2changing carbon monoxide R52 .

(10)  (1   c ) R1  R2 
m

Mc
R3 ,
M1

M1
' R52   g (1   с ) R1  R5 ,R53  0,
2M 2
EE
R1  k1 11 exp(  1 ), R2  k 2  2 2T 0.5 exp(  2 ),RTS
RTS
R51   R5 

(11)

R3  k3 3 S c1 exp( 

666

E3
),
RTS

GEOMATE- Bangkok, Nov. 14-16, 2016

c M 
R5  k5 M 2  1 
 M1 

0.5

evaporated. Then decomposition process into condensed andvolatile pyrolysis products starts, the latter being ignited in
the forest canopy.

 c2 M  2.25
E
T
exp(  5 ) .
M
RT
2 


The initial values for volume of fractions of condensed
phases are determined using the expressions: d (1  z )Wd,  2e , 3e  c 1e 1 .
1e 
1
2
3
where d -bulk density of forest combustible materials, z –coefficient of ashes of forest fuel, W – forest fuel moisture
content. It is supposed that the optical properties of a
medium are independent of radiation wavelength (theassumption that the medium is “grey”), and the so-called
diffusion approximation for radiation flux density were used
for a mathematical description of radiation transport during
forest fires. To close the system (1)–(7), the components ofthe tensor of turbulent stresses, and the turbulent heat andmass fluxes are determined using the local-equilibrium
model of turbulence (Grishin, [1 processes of transfer within
the entire region of the forest massif, which includes thespace between the underlying surface and the base of theforest canopy, the forest canopy and the space above it,while the appropriate components of the data base are used
to calculate the specific properties of the various forest
strata and the near-ground layer of atmosphere. Thisapproach substantially simplifies the technology of solving
problems of predicting the state of the medium in the firezone numerically. The thermodynamic, thermophysical andstructural characteristics correspond to the forest fuels in the
canopy of a different (for example pine [10,12]) type offorest.

Fig. 2 Temperature of gas (1 - T ) and solid phase (2 - TS );T =T/Te , TS =TS/Te, Te=300K.

Fig. 3 Concentration of oxygen (1- C 1 ) and gas products ofpyrolysis (2- C2 ); c  c / c1e , с1e  0.23.

NUMERICAL SOLUTION AND REZULTS
The boundary-value problem (1)–(7) we solve numericallyusing the method of splitting according to physical processes[12]. In the first stage, the hydrodynamic pattern of flow anddistribution of scalar functions was calculated. The system ofordinary differential equations of chemical kinetics obtained
as a result of splitting [12] was then integrated. A discrete
analog was obtained by means of the control volume method
using the SIMPLE like algorithm [12,16]. The accuracy ofthe program was checked by the method of inserted analyticalsolutions. The time step was selected automatically. Fields oftemperature, velocity, component mass fractions, and volume
fractions of phases were obtained numerically. Figure 2
illustrate the time dependence of dimensionless temperaturesof gas (1) and condensed phases (2), Figure 3. – massconcentrations of gas components (1- oxygen, 2- gas productsof pyrolysis), and Figure.4 - relative volume fractions of solid
phases (1), moisture (2) and coke (3) at crown base of theforest (Ve=5m/s). At the moment of ignition, the gas
combustible products of pyrolysis burn away, and theconcentration of oxygen is rapidly reduced. The temperaturesof both phases reach a maximum value at the point of ignition.
The ignition processes are of a gas - phase nature, i.e. initially
heating of solid and gaseous phases occurs, moisture is667

Fig. 4 Volume fractions of solid phase:1  1  1 / 1e , 2   2   2 /  2e , 3   3   3 /  3e .

The vector field of velocities and isotherms of gas phase(Fig. 5): 1-6 correspond to the isotherms T = 1.1, 1.5, 1.7,
2, 3 and 4. In the vicinity of the source of heat and mass
release, heated air masses and products of pyrolysis andcombustion float up. The wind field in the forest canopy
interacts with the gas-jet obstacle that forms from thesurface forest fire source and from the ignited forest canopy
base. Recirculating flow forms beyond the zone of heat andmass release, while on the windward side the movement of
the air flowing past the ignition region accelerates. Under
the influence of the wind the tilt angle of the flame isincreased. As a result, this part of the forest canopy, whichis shifted in the direction of the wind from the center of thesurface forest fire source, is subjected to a more intensive

GEOMATE- Bangkok, Nov. 14-16, 2016

warming up. The isotherms of the gas and condensed
phases are deformed in the direction of the wind.

spread in the forest canopy. The distribution of temperature,
concentrations of gas products of pyrolysis and oxygen in
the forest fire front are presented in the Figure 8. It is seenthat the combustion wave looks like as a soliton.

Fig. 5. The vector field of velocities and isotherms of gas
phase; (t=4 sec.).
Figure 6 presents isolines of oxygen for different instants oftime (t=2, 3, 4 sec.): 1-5 correspond to the isolines c1 = 0.9,
0.8, 0.7, 0.5 and 0.1. Figure 7 presents isolines of gas
products of pyrolysis: 1-4 correspond to the isolines c 2 =0.02, 0.05, 0.1 and 0.5.

Fig. 8 The distribution of gas temperature, concentrations ofgas products of pyrolysis and oxygen.
The oxygen concentration drops to near zero in front of afire. It is consumed in the combustion of the pyrolysis
products, the concentration of which is reached theirmaximum before the maximum of temperature. Figures 9 10
present
thedistributionoftemperature
T (T  T / Te ,Te  300K ) (1- 1.5, 2 - 2., 3 – 2.6, 4 – 3, 5 –3.5, 6 – 4.) for gas phase, concentrations of oxygen c1 (1 –0.1, 2 – 0.5, 3 – 0.6, 4 – 0.7, 5 – 0.8, 6 – 0.9) and volatile
combustible products of pyrolysis c 2 (1 – 1., 2- 0.1, 3 –0.05, 4 – 0.01) ( c  с / c1e , с1e  0.23 ) at different
instants of time for wind velocity Ve= 5 m/s (Fig.9) and Ve=
10 m/s (Fig.10). The distribution of isotherms of
combustion temperature shows the moving of forest fire
front with time. Figure 10 shows that with the increase ofwind speed up to 10 m/s increases the rate of fire spread to 5
m/sec. Ignition of forest due to spotting is one of the mostdifficult aspects to understand the behavior of fires. The
phenomenon of spotting fires comprises three sequential
mechanisms: generation, transport and ignition of recipient
fuel.

Fig. 6 Isolines of oxygen (t= 4 sec.).

Fig. 7. Isolines of gas products of pyrolysis (t=4 sec.).
At Ve  0 the forest fire source ignites the forest crownand burn away in the forest canopy. Forest fire begins to668

GEOMATE- Bangkok, Nov. 14-16, 2016

Fig. 9 The distribution of a) temperature for gas phase, b) concentration of oxygen and c) volatile combustible productsof pyrolysis; Ve= 5 m/s, at different instants of time: I - t=3 sec., II - t=6 sec, III - t=12 sec., IV - t= 20 sec.

Fig. 10The distribution of a) temperature for gas phase, b) concentrations of oxygen and c) volatile combustible productsof pyrolysis; Ve= 10 m/s, at different instants of time: I - t=3 sec., II - t=6 sec, III - t=12 sec., IV - t= 20 sec.

Fig. 11. The distribution of temperature, concentrations of oxygen and volatile combustible products of pyrolysis.

Fig. 12. The distribution of temperature, concentrations of oxygen and volatile combustible products of pyrolysis.
The present mathematical model and results of calculation
are used to illustrate picture of the formation of large fires
by combining small combustion sources arising from thetransfer of firebrands. In order to understand thesemechanisms, many calculation experiments have been
performed. In the Figures 11 - 12 the process of formation
large forest fire front as a result of integration of various
sources of combustion is showed. The distributions oftemperature for gas phase(a), concentrations of oxygen (b)and volatile combustible products of pyrolysis (c) atdifferent times (I - t=3 sec., II - t=6 sec, III - t=12 sec.) for
Ve= 5 m/s are presented. There are present the same valuesof isotherms and isolines of concentration for and as wellas in the Figures 11-12. If the sources of ignition from theburning particles are arranged in a triangle, the processes offormation of the forest fire front are shown in Figures 11-12.

In the second case (Figure 12), the right ignition source inthe x2 direction was 2 times less than in the first case (Figure
11).
CONCLUSION
Results of calculation give an opportunity to describe thedifferent conditions of the forest fire spread taking accountdifferent weather conditions and state of forest combustible
materials, which allows applying the given model for
prediction and preventing fires. It overestimates the rate ofcrown forest fire spread that depends on crown properties:
bulk density, moisture content of forest fuel, wind velocity
and etc. The model proposed here gives a detailed picture of
the change in the temperature and component concentration
fields with time, and determine as well as the influence ofdifferent conditions on the crown forest fire spreading forthe different cases of inhomogeneous of distribution ofsources of forest fires initiation.

669

GEOMATE- Bangkok, Nov. 14-16, 2016
ACKNOWLEDGEMENTS
This research was financially supported by the RussianFoundation for Basic Research in 2016 (the project № 1641-700022 р_а) and on the state task № 2014/64, the state
project “Scientific researches organization”.
REFERENCES[1] Van Wagner CE, “Conditions for the start and spread ofcrown fire”, Canadian Journal of Forest Research, vol.
7, 1977, pp. 23-34.
[2] Alexander ME, “Crown fire thresholds in exotic pine
plantations of Australasia”, PhD thesis, Department ofForestry, Australian National University, Australia,
1998.
[3] Van Wagner CE, “Prediction of crown fire behavior inconifer stands”, Proc. '10th conference on fire andforest meteorology. Ottawa, Ontario. (Eds D. C.
MacIver, H. Auld and R. Whitewood), 1989, pp. 207212.
[4] Xanthopoulos G., “Development of a wildland crownfire initiation model”, PhD thesis, University ofMontana, USA, 1990.
[5] Rothermel RC, “Crown fire analysis and interpretation”,
Proc. International conference on fire and forest
meteorology. Missoula, Montana, USA, 1991.
[6] Rothermel RC, “Predicting behavior of the 1988
Yellowstone Fires: projections versus reality”, Int.
Journal of Wildland Fire, vol. 1, 1991, pp. 1-10.
[7] Van Wagner CE, “Prediction of crown fire behavior intwo stands of jack pine”, Canadian Journal of Forest
Research, vol. 23, 1999, pp. 445-449.

670[8] Cruz MG, “Predicting crown fire behavior to support
forest fire management decision-making”, Proc. IV
International conference on forest fire research, LusoCoimbra, Portugal. (Ed. D. X. Viegas), 11 [CD-ROM].
(Millpress), 2002.
[9] Albini FA, “Modeling ignition and burning rate of large
woody natural fuels”, Int. Journal of Wildland fire, vol.
5, 1995, pp. 81-91.
[10] Scott JH, “Assessing crown fire potential by linking
models of surface and crown fire behavior”, USDA
Forest Service, Rocky Mountain Forest and Range
Experiment Station. Fort Collins: RMRS-RP-29,(Colorado, USA), 2001.
[11] A.M. Grishin, Mathematical Modeling Forest Fire and
New Methods Fighting Them, Publishing House of
Tomsk University, Tomsk, Russia, 1997.
[12] Grishin AM, Perminov VA, “Mathematical modeling of
the ignition of tree crowns”, Combustion, Explosion,and Shock Waves, vol. 34, 1998, pp. 378-386.
[13] Konev EV, The physical foundation of vegetative
materials combustion, Nauka, Novosibirsk, Russia,
1977.
[14] Morvan D, Dupuy JL, “Modeling of fire spread througha forest fuel bed using a multiphase formulation”,
Combustion and Flame, vol. 127, 2001, pp. 1981-1994.
[15] Morvan D, Dupuy JL, “Modeling the propagation of
wildfire through a Mediterranean shrub using a
multiphase formulation”, Combustion and Flame, vol.
138, 2004, pp. 199-210.
[16] Patankar SV, Numerical Heat Transfer and Fluid Flow,
Hemisphere Publishing Corporation, New York, 1981.

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

UNIT WEIGHT AND COMPRESSIVE STRENGTH OF CELLULAR
LIGHTWEIGHT RECYCLED GLASS-FLY ASH GEOPOLYMER

Cherdsak Suksiripattanapong1, Rutchakit Maythathirut2, Sermsak Tiyusangthong3, Suksun Horpibulsuk4
and Arul Arulrajah5
1,3School of Civil Engineering, Faculty of Engineering and Architecture, Rajamangala University ofTechnology Isan, Thailand
2,4
School of Civil Engineering, and Center of Excellence in Innovation for Sustainable Infrastructure
Development, Suranaree University of Technology, Thailand
5
Faculty of Science, Engineering and Technology, Swinburne University of Technology, Australia

ABSTRACTThis paper investigates unit weight and compressive strength of cellular lightweight recycled glass-fly ash
geopolymer. The Recycled Glass (RG) was collected from Gaew Grung Thai Co., Ltd. in Thailand and Fly Ash(FA) was obtained from the Mae Moh power plant of the Electricity Generating Authority of Thailand. The
influential factors studied are liquid alkaline activator content (L), NaOH concentration and air content (A c ).
Test results show that unit weight of cellular lightweight RG-FA geopolymer decreases significantly as the A c
increases and regardless of L and NaOH concentration. The 7-day compressive strength of cellular lightweight
RG-FA geopolymer increases with increasing NaOH concentration. However, the excessive of L and A c resultsin the reduction of 7-day compressive strength of RG-FA geopolymer. The maximum 7-day compressive
strength of RG-FA geopolymer is found at L/FA ratio of 0.6, Na 2 SiO 3 :NaOH ratio of 70:30 and NaOH
concentration of 7 M, which is 20, 14, 5 and 2.5 MPa for air content of 0, 1, 3 and 5%, respectively.
Keywords: Cellular Lightweight, Recycled Glass, Geopolymer, Fly Ash

Thailand [12]Suksiripattanapong et al. [13] investigatedstrength development in water treatment sludge-FA
geopolymer. The optimum ingredient providing
maximum
unit
weightandstrength
is
NaOH/Na 2 SiO 3 ratio of 80:20 and L/FA ratio of 1.3,
irrespective of heat condition and curing time. The
optimum heat temperature and duration for the
optimum ingredient are 75oC and 72 hours,
respectively. The durability against wet-dry ofsludge-FA geoplymer is better than that of claycement [14].
Recently, Arulrajah et al. [8] investigatedstrength and microstructure evaluation of RG-FA
geopolymer as low-carbon masonry units. The
Na 2 SiO 3 /NaOH ratio of 70/30 with L/FA ratio of
0.625 were most efficient combination to provide
the required unconfined compression strengths for
manufacturing masonry. The maximum strength of
RG-FA geopolymer was about 17 MPa.
This research investigates unit weight andcompressive strength of cellular lightweight RG-FA
geopolymer. The influential factors studied are
liquid alkaline activator content (L), NaOH
concentration and air content (A c ). This researchenables RG traditionally destined for landfill to be
used in a sustainable manner in cellular lightweight

INTRODUCTION
Massive amounts of waste glass are produced
worldwide every year. In Australia, approximately
1.0 million tonnes of Recycled Glass (RG) is
destined for landfill annually [1]. In the UK,
households throw away over 29.1 million tonnes ofwaste annually [2]. RG is the by-product of crushingmixed color bottles and other glass products
collected from Gaew Grung Thai Co., Ltd. in
Thailand. RG is mainly composed of sand and silt
particles with a small percentage. RG has recently
been used in construction materials such as concrete[2,3], roads [4-7], footpaths [8] and masonry unit [1].
Fly Ash (FA) derived from coal-fired electricity
generation provides the greatest opportunity forcommercial utilization of this technology due to theplentiful worldwide raw material supply [9]. Palomo
et al [10] found that the different FA activated with
8-12 M NaOH cured at 85oC for 24 hours produced
a material with compressive strength of 35-40 MPa
and about 90 MPa if sodium silicate (Na 2 SiO 3 ) is
added to the NaOH solution. Xie & Yunping [11]reported that the hardening process of FA activatedwith Na 2 SiO 3 is mainly attributed to the gel-likereaction products that bind FA particles together. FA
is extensively used as a precursor for geopolymers in
671

GEOMATE- Bangkok, Nov. 14-16, 2016

masonry unit.
MATERIALS AND METHODS
Materials
Recycled glass (RG) was collected from Gaew
Grung Thai Co., Ltd. in Thailand. Figure 1 indicatesthe particle size of RG determined by laser particle
size analysis, indicating that an average grain size,
D 50 , of RG is 0.592 mm and a specific gravity is
2.62. Table 1 shows the chemical compositions of
RG. It is composed of high SiO 2 content of 85.64%
and the CaO content is 10.63%.
FA was obtained from the Mae Moh power plantof the Electricity Generating Authority of Thailand(EGAT) in the northern region of Thailand. Table 1
summarizes the chemical composition of FA using
X-ray ﬂuorescence (XRF). Total amount of the
major components (SiO 2 , Al 2 O 3 and Fe 2 O 3 ) are
77.15% while the CaO content is 16.09%. Figure 1
shows the grain size distribution curve of FA, which
was tested by laser particle size analysis. It is shownthat the RG particles are larger than the FA ones.
The average grain size of FA is 17.37 micron. The
specific gravity of FA is 2.31. The morphology of
the RG and the FA is shown in Figure 2. The FA
particles are fine and spherical whereas the RG
particles are irregular in shape. The liquid alkaline
activator is a mixture of Na 2 SiO 3 , which consists of
9% Na 2 O and 30% SiO 2 by weight, and NaOH withvarious concentrations of 1, 3, 5 and 7 molars.
Air foam agent, Sika Poro 40 from Sika
(Thailand) Company limited, was used in this study.
Sika Poro 40 is a blend of anionic surfactants andfoam stabilizers. It is a liquid air entraining agent
used in various types of mortar, concrete andcementitious material. The air foam was prepared bymixing the foaming agent with water at a ratio of
1:50 by weight.

RG

FA
Fig. 2

Percent finer (%)(SEM)

Table 1 Chemical composition of recycled glass andﬂy ash.
ChemicalRecycled
Fly ashcomposition (%)
glass
SiO 2
85.64
43.10
Al 2 O 3
N.D.
18.95
Fe 2 O 3
0.58
15.10
CaO
10.63
16.09
MgO
2.81
N.D.
SO 3
N.D.
3.66
Na 2 O
N.D.
0.73
K2O
0.16
1.76
LOI
0.18
0.61

100
80

Scanning electron microscopy
images of RG and FA.

FA
Recycled glass

Remark: N.D. = not detected

60

Sample Preparation

40

The cellular lightweight RG-FA geopolymer
sample is a combination of RG, FA, liquid alkaline
activator (Na 2 SiO 3 and NaOH) and air content (A c ).
The RG/FA ratio was fixed at 50:50. The
Na 2 SiO 3 /NaOH ratio was 70:30. The FA and L weremixed for 5 minutes in a mixer to ensure
homogeneity of the mixture. The input L contents inthe cellular lightweight RG-FA vary from 0.6 to 0.8
times FA. The mixer was stopped and then mixed
with RG for additional 5 minutes. The air form wasadded into the RG-FA-L mixture and mixed for 5

20
0
0.1

Fig. 11

10100Particle Diameter (µm)

1000

10000

Grain size distribution of RG and FA.

672

GEOMATE- Bangkok, Nov. 14-16, 2016

minutes. The A c values were 1, 3 and 5% by total
weight of FA. The uniform cellular lightweight RGFA geopolymer was transferred to containers of
50x50x50 mm. The samples were dismantled,wrapped within vinyl sheet and subsequently cured
at room temperature (27-30οC). Unit weight andstrengths of cellular lightweight RG-FA geopolymer
samples were measured after 7 days of curing inaccordance with ASTM C138 and ASTM C69-09,
respectively.

RG-FA geopolymer at different air content.

UNIT
WEIGHTANDCOMPRESSIVE
STRENGTH OF CELLULAR LIGHTWEIGHTRECYCLED GLASS-FLY ASH
Figure 3 shows the relationships between 7-day
unit weight versus concentration of NaOH ofcellular lightweight RG-FA geopolymer at different
air content for Na 2 SiO 3 /NaOH ratio of 70:30. The
amount of liquid alkaline activator is 0.6FA, 0.7FA,
and 0.8FA. The air content (Ac) values varied
between 1 and 5% by weight of fly ash. The test
result shows that the unit weight of cellular
lightweight RG-FA geopolymer decreases as aircontent and L increases. The unit weight of cellular
lightweight RG-FA geopolymer increases asconcentration of NaOH increases. It is because thedensity of the NaOH solution increases with theincrease in concentration of NaOH.
Fig. 4

Relationship between compressive strengthand NaOH concentration of cellular
lightweight RG-FA geopolymer at different
air content.

Figure 4 shows the relationship between 7-daycompressive strength and NaOH concentration ofcellular lightweight RG-FA geopolymer at different
A c . The amount of L was 0.6FA, 0.7FA and 0.8FA
at Na 2 SiO 3 /NaOH ratio of 70:30 and A c of 0, 1, 3
and 5%. The test result shows that the 7-daycompressive strength of cellular lightweight RG-FA
geopolymer varies with the L content for all A c . The
7-day compressive strength of cellular lightweight
RG-FA geopolymer decreases with increasing L. It
is because the excessive liquid alkaline activator
results in the bleeding in sample [15].
The
maximum compressive strength of cellular
lightweight RG-FA geopolymer is at L = 0.7FA for
all A c . Compressive strengths of cellular lightweight
RG-FA geopolymer on L = 0.6FA and L = 0.8FA
are similar.
CONCLUSION

Fig. 3The unit weight and compressive strength ofcellular lightweight Recycled Glass-Fly Ash (RGFA) geopolymer are evaluated in this study. The
conclusions can be drawn as follows.
1. The unit weight of cellular lightweight RG-

Relationship between unit weight andNaOH concentration of cellular lightweight673

GEOMATE- Bangkok, Nov. 14-16, 2016

FA geopolymer increases as NaOH concentration
increases. It is because the density of the NaOH
solution increases with the increase in NaOH
concentration.
2. The 7-day compressive strength of cellular
lightweight RG-FA geopolymer decreases withincreasing L. It is because the excessive liquidalkaline activator results in the bleeding in sample
3. The maximum compressive strength ofcellular lightweight RG-FA geopolymer is at L =0.7FA for all air contents. Compressive strengths ofcellular lightweight RG-FA geopolymer at L =0.6FA and L = 0.8FA are similar.

[6] Ooi P., Li M., Sagario M., Song Y. Shearstrength characteristics of recycled glass,
Transp. Res. Rec.: J. Transp. Res. Board 2059
(2008) 52–62.
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environmental impacts of recycled glass used asa pavement base material, Waste Manage. Res.
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[8] Arulrajah A., Ali M., Disfani M., PiratheepanJ., Bo M. Geotechnical performance of recycled
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[13] SuksiripattanapongC.,
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geopolymer masonsy units manufactured fromwater treatment sludge”, Construction and
Building Materials, 2015, Vol.82, pp.20-30.
[14] HorpibulsukS.,
Suksiripattanapong
C.,
Samingthong W., Rachan R. & Arulrajah A.,“Durability against wet-dry cycles of water
treatment sludge-fly ash geopolymer”, Journalof Materials in Civil Engineeing (2015)[15] Posi, P., Teerachanwit, C., Tanutong, C.,
Limkamoltip, S., Lertnimoolchai, S., Sata, V.,
Chindaprasirt, P. “Lightweight geopolymer
concrete containing aggregate from recycle
lightweight block.” Materials and Design
Vol.52, 2013, pp.580–586.

ACKNOWLEDGEMENTS
This work was financially supported by theThailand Research Fund under the TRF SeniorResearch Scholar program Grant No. RTA5680002.
The financial support from the Office of Higher
Education Commission under NRU project ofThailand, as well as Suranaree University ofTechnology are also very much appreciated. The
first author acknowledges the financial support fromthe Rajamangala University of Technology Isan.
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S. Recycled-glass blends in pavement
base/subbase applications: laboratory and field
evaluation, J. Mater. Civ. Eng. 26 (7) (2013)[5] Grubb D., Gallagher P., Wartman J., Liu Y.,
Carnivale M. Laboratory evaluation of crushed
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FINITE ELEMENT SIMULATION OF EMBANKMENTS UTILIZINGSMALL-SCALE CENTRIFUGAL DIMENSIONSAli Sobhanmanesh1,2, Ramli Nazir3, Tan Choy Soon1 and Deprizon Syamsunur1
Department of Civil Engineering, Faculty of Engineering, Technology and Built Environment, UCSI University,
56000, Kuala Lumpur, Malaysia;
2
Process System Engineering Center, Faculty of Engineering, Technology and Built Environment, UCSI
University, 56000, Kuala Lumpur, Malaysia
3
Department of Geotechnics & Transportation, Faculty of Civil Engineering, University Teknologi Malaysia (UTM),
Malaysia
1

ABSTRACT
Four different cases of unreinforced and reinforced embankment models constructed on soft and stiff
grounds were studied. Small-scale physical modelling by means of centrifuge tests and numerical modelling bymeans of finite element simulations were performed. As the small-scale model was rotated in different
acceleration fields during the centrifuge test, the dimensions of the centrifugal model were different from theoriginal states of the prototype in different stages of the test. This paper focused on developing a finite element
simulation based on the dimensions of a centrifugal model in different incremental acceleration fields applied
during the stages of the test. Comparing the results of finite element simulations with the measurements of thecentrifuge tests showed a good agreement between the two methods, which verified the reasonableness of thefinite element models in analysis of embankments based on small-scale centrifugal dimensions. Moreover, The
results showed the different deformation behaviour for embankments on soft and stiff grounds and indicated thesignificant effect of the geosyntheic reinforcement on increasing the stability of the embankment on soft ground.
Keywords: Finite Element Simulation, Centrifuge Test, Reinforced Embankment, geosynthetics

There are two methods of the staged-constructionof a model in the centrifuge test: 1) constructing themodel in flight using raining technique 2) applying
incremental acceleration fields in different stages of
construction on a pre-constructed model withconstant dimensions. The first method results inmore accurate results but needs more advanced fully
equipped apparatus. However, due to the limitations,second method is also used commonly in many
centrifuge tests.
A number of these centrifuge tests have beenmodelled by numerical approaches. In numericalsimulation of centrifugal models, most ofresearchers have only utilized the full-scale
prototype dimensions. [1]-[11]. It should be
mentioned that using the second method, acentrifugal model has same dimensions and samestress levels only at the final stage (final acceleration
field) of the test compare to the prototype, i.e. thecentrifugal model has different dimensions andstress level in different acceleration fields before
reaching the final stage. Therefore, a numericalsimulation, which can consider these differences indimensions and stress levels in varying acceleration
fields of the centrifuge test, is very important andcan consequence in more accurate results.
Nevertheless, such simulations are rare and limited
so far.

INTRODUCTION
The behaviour of geosynthetically-reinforced
structures has been studied through observation of
full-scale physical models by many researchers.
However, The cost of construction and monitoring
the full-scale embankment model is quite high and isa time consuming procedure with some limitations.
Hence, an alternative method such as scaled-down
modelling can be used. Although a small-scalemodel is relatively economical, in terms of financeand time, it is not reliable in predicting the actual
prototype behavior due to the differences in stresslevels surrounding the model and prototype
respectively. For this reason the centrifugalmodeling technique has become increasingly used as
it has ability to reproduce the same stress levels in asmall-scale dimension model as those present in afull-scale prototype. Centrifuge modeling also hassome limitations, e.g. the prototype is under earth’s
gravity field where the radius of the earth is infinite
compared to the prototype size. Thus the earth's
gravitational field will act parallel and be uniform in
direction at all points in the prototype. However inthe centrifuge, since it has finite radius compared tothe model size, the artificial gravitational field isradial, which is non-linear and non-uniform.

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Therefore, the focus of this paper is to conduct a
realistic numerical simulation of centrifugal models
utilizing small-scale centrifugal model dimensions to
consider exact procedure of a centrifuge test
utilizing abovementioned factors, i.e. different
dimensions and stress levels in different applying
acceleration fields of the test. Comparing the resultsof theses numerical simulations with measurementsof centrifugal models validates this numericalmodelling technique.
EMBANKMENT MODEL CASESIn this research, four embankment models withdifferent foundation soil and reinforcement
condition were considered in both centrifugal andnumerical modelling. Details of these cases areshown in Table 1. A plan view and cross section ofembankment model used in both centrifuge andnumerical modeling is shown in Fig. 1. The scaleddown embankment model has a height of 5 cm, acrest width of 14 cm and slope of 1V to 1H,
underlain by 7 cm foundation soil. Due to inherent
symmetry about the centerline, only one half of themodel was considered.

CENTRIFUGE TEST
Different centrifuge tests were conducted using a
mini centrifuge apparatus of Universiti Kebangsaan
Malaysia (UKM). It is a beam-type centrifuge,
designed to allow centrifuge testing of soil package
up to 6 kg with a maximum rotational speed of 500
rpm and can accelerate up to 140 g at an effective
radius of 0.5 m [12].
Because of limitations and technical difficultiesof UKM mini-centrifuge apparatus in constructing ashaped embankment in flight, the centrifuge model
was constructed before the test running andincremental accelerations was applied to simulate
staged-construction of centrifugal modeling. In thismethod, the acceleration increased slowly to reachthe predefined level. The rotation was maintained inthis specific acceleration level for a certain time, andthen the acceleration increased to reach the nextlevel. Details of centrifuge test processes are
depicted in Table 2.
Table 2 Centrifuge test proceduresω (rpm)Table 1 Details of four embankment cases
Case

Foundation

Fill material

Reinforcement

I
II
III
IV

KaolinKaolinSand
Sand

Claye sand
Claye sand
Claye sand
Claye sandTextileTextile16 cm 5cm

14 cm 5 cm 16
10 cm

56 cm
16 cm 5cm

14 cm 5 cm 16 cm
5 cm 7 cm
56 cm

Fig. 1

Plan view and cross-section of embankment
model

130
184
225
260
291

Gravity (g)g = (1.18×103) ×r*×(ω) 2
10
20
30
40
50

Maintained
Time (minute)
5
5
5
15
30

FINITE ELEMENT MODELINGIn this research, numerical modeling by means offinite element simulations were conducted using
PLAXIS software, which is a powerful program in
simulating and analyzing the geotechnical problems.
Properties of materials used in the finite element
analysis are shown in Table 3. These results are
adopted from laboratory tests. The undarined shear
strength (Su) of kaolin considered as 8 kPa, which
was defined by a mini-vane shear test.
Table 3 Properties of materials in FE modeling
Parameters
Material Model
Behavior
Unit weight(kN/m3)
PermeabilityΚ x = K y (m/day)
Young’s Modulus(kN/m2)
Poison’s ratio ν
Cohesion(kN/m2)
Friction angle φ

676

ClayeySand
MC
DrainedSand
MCDrained

KaolinMC
Undrained

18

19

16

0.5

1

3 x 10-4

6000

10000

1000

0.3
5
30

0.3
1
31

0.35
8
0

GEOMATE- Bangkok, Nov. 14-16, 2016

The standard fixities were used to define theboundary conditions, i.e. full fixity (rough rigid
boundary) was assumed along the bottom of themodel and the vertical boundaries of the model were
fixed in the horizontal direction (smooth rigid
boundary). Geogrid reinforcement was modeled asline elements with two translational degrees of
freedom in each node (ux, uy). The only material
property of the geogrids in 2-D simulation is anelastic normal (axial) stiffness EA determined fromthe curve of the elongation of the geogrids plottedagainst the applied force and is the ratio of the axial
force per unit width to the axial strain. The interface
elements are considered to model the interaction
between reinforcement and soil. The roughness of
the interaction is modeled by choosing a suitablevalue for the strength reduction factor in theinterface (R inter ). Based on the direct shear tests
carried on the interface of the soil and textile, theinterfaces ratio between geotextile and soil was
calculated as 0.95 for this study.

In PLAXIS, the earth gravitational acceleration(g) can be simulated by considering the multiplier(ΣMweight) equal to 1. Therefore, the process ofvarying incremental centrifugal accelerations was
simulated in PLAXIS by considering different
multiplier of ΣMweight more than 1 related todifferent acceleration fields of centrifuge test (5 g,
10 g, 20 g, 30 g, 40 g, and 50 g) and also correspondto time period of rotation in that acceleration level asshown in Fig.3. Increasing the acceleration up to
five levels of 10 g, 20 g, 30 g, 40 g, and 50 g
simulated a five-staged construction of theembankment in centrifuge test. The five level
accelerations corresponded to the construction
heights of 0.5 m, 1 m, 1.5 m, 2 m, and 2.5 m,
respectively, according to the law of similarity. Aftereach staged construction, a rest period wasconsidered in the centrifuge model that is almost
equal to the 10 hours for thickness of 0.5 m, 1.0 day
for thickness of 1 m, 3 day for thickness of 1.5 m,
16.5 days for thickness of 2.0 m, and 52 days for
thickness of 2.5 m.

Simulation Procedure
Construction stages by means of plastic analysis
and post-construction stages by means ofconsolidation analysis were utilized as PLAXIS
allows for undrained, drained and consolidation
analyses of two-dimensional plane strain or
axisymmetric problems. The elastic perfectly plastic
materials with consideration of Mohr–Coulomb
failure criterion were utilized to model the behaviorof the embankment and foundation materials for allcases; the reinforcement element was modeled aslinear-elastic material model. Fig. 2 shows thegenerated mesh and the considered plate at
centerline of the embankment to simulate theinteraction between soil and side-wall of strong box.

Rest period related toconsolidation analysis after
each stages of construction
Actual sequence of staged-construction

Fig. 3

Sequence staged-construction of modelwith varying acceleration fields

RESULTS AND DISCUSSION
Considered plate at centreline of the embankment

Fig. 2

generated mesh and boundary fixities

To consider varying dimensions of model due tothe different acceleration fields of centrifugal tests,
the FE modelling was performed based on the smallscale centrifugal model dimensions but under theaccelerations more than earth gravitational
acceleration (g).

The results of centrifuge tests and FE analyses
for all cases are described in the following. A sideby side comparison between the FE results andcentrifuge measurements are also depicted at the lastsection. Deformed mesh and shading of verticaldisplacements for different cases are shown in Figs.
4, 5 and 6. It is clear that vertical and horizontal
displacements occurred at the embankment fill,
beneath the fill and at toe of the embankment. For
cases constructed on soft ground i.e. cases I & II(refer to Table 1), slight heave and uplift occurred at
the embankment toe and ground surface beyond thetoe due to the undrained behaviour of subsoil layers.
The extreme total displacements occurred at topsideof the fill slope at embankment crest. The maximum
displacement of case II at topside of fill slope ofembankment computed as 3.1 mm, which reduced
about 24.4% compare to the displacements of the677

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unreinforced model (4.1 mm for case I). The upliftof the embankment toe was also decreased
significantly compare to case I. This shows theeffect of the geotextile reinforcement on increasing

the stability and reducing the deformations of theembankment constructed on soft ground.

Toe Uplift and soil heave

Fig. 4(a) Deformed mesh (b) vertical displacements (case I)

Fig. 5

Deformed mesh (b) vertical displacements (Case II)

Fig. 6

Deformed mesh (b) vertical displacements (Case III)For cases constructed on stiff ground (cases III &IV), there was not any considerable difference indeformations and displacements of unreinforced andreinforced models. It indicated that the geosynthetic
reinforcement have small influence on deformation
behaviour of embankments on stiff ground becauseof small induced vertical displacements and lateral
movements.
Comparison Between the Results of the Finite
Element and Centrifuge Modelling

A side-by-side comparisons between the
deformation
patterns
resultedfromthemeasurements of centrifuge tests and analysis offinite element simulations for four different cases areshown in Fig. 7. It is clear that the deformation
patterns of the centrifugal model test and finite
element simulation are almost similar for all cases,
which validates the simulation of centrifugal model
tests, using finite element simulation, based on
small-scale modelling procedure.

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Fig. 7

Side-by side comparison of the deformation patterns resulted from centrifuge and FE models a) Case I,b) Case II, c) Case III, d) case IV

A comparison between resulted verticaldisplacements of the centrifuge test and FE analysis
for all cases are shown in Fig. 8. Moreover, themaximum vertical displacement obtained from
centrifuge test and FE analysis is presented in Table
4. The coincidence of the lines correspond to
centrifuge tests and FE analysis for each case in Fig.
8 and results of Table 4 indicates a good agreement
between displacements obtained from these two
methods of modelling.
CONCLUSIONS AND SUMMARY
Small-scale physical modeling and numericalanalysis of different embankment cases have been
carried out using centrifugal test and finite element
simulations. Important affective factors have beenstudied to provide a better understanding and deeper
insight of the embankment behavior under different
conditions. In reinforced embankments, the tension

mobilized in the reinforcement can be very effectivein improving the short-term stability of theembankment. The tension in the reinforcement issensitive to the magnitude and distribution of shear
strength of clay foundation. Basal reinforcement of
the embankment using geotextile resulted in
reduction of the displacements up to 25% and leads
to construction of higher fill embankment or steeper
slope. Under a relatively small deformation,
however, the influence of interfaces behavior ofgeosynthetic reinforcement and soil on the system
performance is expected to be less important as
observed in cases on stiff ground (Cases III and IV).
Finite element simulations are successful in
predicting the overall behavior of unreinforced andreinforced embankments. The results of FE
simulation using full-scale prototype dimensions canhave some differences with centrifuge measurements.
However, FE simulations using small-scalecentrifuge dimensions and actual conditions of test

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with considering affective parameters (e.g. varying
gravitational acceleration level, side-wall friction
and small boundary condition), minimizes thedifferences between FE predictions and centrifugal
measurements. It is shown a good agreement
between results of these two methods, which
validates the finite element simulation.

0

0

20

Time (min)40

Settlement (mm)

-0.5
-1
-1.5
-2
-2.5

centrifuge measurement case I
centrifuge measuement Case II
centrifuge measurement Case III
FE case I
FE case II
FE case III

-3
-3.5
-4
-4.5

Fig. 8vertical displacement of different cases
from centrifuge test and FE analysis

Table 4 Maximum settlements resulted from
centrifuge test and FE analysis
Cases
Centrifuge Test
FE analysis
Case I
4 mm
4.10 mm
Case II
3 mm
3.10 mm
Case III1 mm
1.0 mmCase IV
1 mm
1.0 mmThe effect of side friction between soil andsidewall on deformation behavior can be significant
on the deformation pattern and behavior of models.
Efforts should be made in the centrifuge work tominimize the effects of side friction by using somematerials like silicon grease to reduce the friction.
ACKNOWLEDGMENT
The first author would like to thank the faculty of
Science and Technology, Universiti Kebangsaan
Malaysia (UKM) for providing centrifugal facilities.
REFERENCES[1] Ding, J. H. and Bao, C. G., “Centrifugal model
test and finite element analysis of geosyntheticreinforced embankments on soft ground anddredger fill”, China Civil Eng Journal, 1999.
[2] Sharma, J. and Bolton, M. D., “Centrifugal andnumerical
modellingof
reinforced
embankments on soft clay installed with wick60

drains”, Geotextiles and Geomembranes, 19 (1),2001, pp. 23–44.
[3] Hu, H. R., Chen, S. L. and Shen, Z. L.,“Centrifugal & numerical modeling ofgeosynthetic reinforced embankments on softclay”, Rock and Soil Mechanics, 24 (3), 2003,
pp. 389–394.
[4] Wang, H. and Huang, X, “Centrifuge Model
Test and Numerical Analysis of Embankment
Widening on Soft Ground”, ASCE. In: EighthInternational Conference on Applications ofAdvanced Technologies in Transportation
Engineering (AATTE), 2004, Beijing, pp. 548–
553.
[5] Jiang, X., Qiu, Y. J. and Wei, Y. X.,“Engineering
behaviorofsubgradeembankments on sloped weak ground based onstrength reduction FEM”, Chinese Journal of
Geotechnical Engineering, 29 (4), 2007, pp.
618–625.
[6] Chen, J.-F. and Yu, S.-B., “Centrifugal andNumerical Modeling of a Reinforced LimeStabilized Soil Embankment on Soft Clay with
Wick Drains”, International Journal of
Geomechanics, 11 (3), 2011, pp. 167–173.
[7] Zhang, J., Zheng, J. and Huang, X., “Centrifuge
Model Test and Numerical Analysis of
Geosynthetic-Reinforced Embankments for a
Road Widening”, Road Pavement and Material
Characterization, Modeling, and Maintenance,
ASCE, (213), 2011, pp. 204–210.
[8] Chen, J.-F. and Yu, S.-B., “Centrifugal andNumerical Modeling of a Reinforced LimeStabilized Soil Embankment on Soft Clay with
Wick Drains”, International Journal of
Geomechanics, 11 (3), 2011, pp. 167–173.
[9] Li, B., Yu, W., Gong, B. and Cheng, Z.,“Centrifugal and numerical modeling of highand steep geosynthetic-reinforced slopes”. In
Proc. Geo-Congress 2013, ASCE, 2013, pp.
482–488.
[10] Mohamed, S. B., Yang, K. H. and Hung, W.Y.,“Finite element analyses of two-tiergeosynthetic-reinforced soil walls: Comparison
involving centrifuge tests and limit equilibrium
results”, Computers and Geotechnics, 61,
Elsevier Ltd, 2014, pp. 67–84.
[11] Zhang, J., Cen, G., Zhou, C. and Ouyang, Y.,“Centrifuge Model Test and Numerical
Analysis of Properties of a Soft Foundation
with an Upper Crust”, The Electronic Journalof Geotechnical Engineering (EJGE), (19)2014, pp. 17195–17208.
[12] Sobhanmanesh, A., Ramli, N. and Gofar, N.,“3D Numerical Analysis of centrifuge tests onembankments on soft and stiff ground”,
Advanced Materials Research”, (831), 2014, pp.
314-320.

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BEHAVIOR OF ACCIDENTAL CASES OF TBM SEGMENTAL LINING
Mostafa Zaki1, Ahmed Hassan2, and Mostafa Asaad3
1,2

Faculty of Engineering, Minia University, Egypt; 3 NAT, Egypt

ABSTRACTThis research aims to study the behavior of segmental lining at accidental cases. To achieve this goal, first a shortstudy is performed to develop an appropriate model for simulating segmental lining components. Then acomprehensive study is introduced for the effect of parameters which may lead to excavation failure. Investigation
includes deformation and plastic zones generation, around tunnel segments and connecting bolts for different cases
during construction. Consequently better understanding is possible for the interaction between different segmental
lining components and the annular grout. It is concluded that installation of bolts is very important during theconstruction of lining and should be carefully handled. If more than one bolt isn’t installed during the construction of
the ring, ring failure may happen. Excessive grout concentrated at some rings has destructive effect on the ring andmay lead to collapse of ring. Consequently this situation should be totally avoided during the construction of TBM
segmental lining.
Keywords: Tunnels excavation; lining components; safety limits; accidental cases
grout concentrated in certain segments on thestability of the model.

INTRODUCTION
Mechanized tunnelling has speedily spreadworldwide for tunnels excavation due to high safety
degree and high speed rate of work. If an accidental
case happens during construction, the consequences
for time, cost, and human losses are significantly
high. To reduce this risk, the interaction between
lining components with grout and surrounding soil
should be comprehensively understood. The aim ofstudying the nonlinear behavior of TBM segmental
lining is to predict its behavior in accidental cases.
Generally the structural design model should yield
criteria related to failure cases, against which thetunnel should be designed safely [6].
This paper focuses on studying the behavior ofsegmental lining to avoid any possible failure of thelining. Furthermore if the previous goal wouldn’t bepossible in some cases, at least it will be possible toreduce the time of the delay consumed to investigate
the reasons of failure and to overcome them. Firstly asuggested technique is developed to simulate the
TBM segmental lining. In this technique theimportant components of TBM segmental lining(segments, joints, and bolts) are simulated by theirown properties. Two possible accidental cases for
failure are studied in this study. First case focuses on
studying the effect of not installing some or all boltson the stability of the ring during construction. The
second case is concerned with effect of excessive

SOIL MODELLING
During the construction of lining and before thehardening of the grout, there is no direct interaction
between the lining and the soil. This case isconsidered critical during tunnel construction as
most failure cases happen during it. Hence, it isreasonable to model soil confinement by series ofsprings. Simulation of soil as springs will help tomore focus on the local behavior of the lining during
construction.
Stiffness of Springs
The springs’ stiffness was calculated using
linear load deformation relationship according to
Duddeck and Erdman [1], [2], [7] and [8].
(1)
A

(2)
(3)

Where:
E, ν = Young’s Modulus and Poisson’s ratio of soil,
R = equivalent tunnel radius,

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A = area of soil to be represented by the equivalentradial spring. This is distance between adjacent radialsprings multiplied by unit length.
Spring constants is determined based on
tributary projections on the X and Y axis of eachjoint [12]. It worth mentioning that ground behavior
in tangential direction is generally ignored in beamspring model method. Although this leads tofrictionless sliding of the lining against the ground, it
has a negligible effect on member forces [3].

Model Parameters
Data for the model considered in this research istaken from the accident of Cairo metro Line 3 Phase
1, at tunnel section between Bab El-Sharia and ElGiesh stations [10]. Fig. 3 shows the zone of failure
in red in the clouded area. diameter is 8.0 m eightsegments of thickness 0.4 m and length 1.5 m wereused. The used bolts 300 m length and 28 mm
diameter. Model parameters are shown in Table 1.

FEM MODEL
A three-dimensional FEM model was used toperform the analyses. The FEM software package
Midas GTS [9] was used in this research. Fig. 1
shows segmental lining simulation. Segments joints
simulation is shown in Fig. 2. Since the purpose ofthis paper is to study the local behavior of segmental
lining, 3-D simulation will be used. Consequently it
will be possible to accurately simulate and study the
behavior of each component of TBM segmental
lining. In this approach, all elements of segmental
lining are modelled with solid elements
Fig. 2 Simulation of segments joints.

Fig. 1 Numerical simulation of segmental lining.
Fig. 3 Map showing the route of Cairo Metro Line 3.

Table 1 Parameters of different materials used in the analysis
Parameter
Initial Tangent Elastic Modulus (E) MPa
Poisson’s Ratio (ν)
Density kg/m3
Unconfined Compressive Strength( fcu) MPa
Initial Yield Stress (σy) MPa
Tensile failure stress MPa
Type of Selected Element
Model Behavior

Segments (Concrete)
30000
0.2
2600
60
48
4
Solid Element
Drucker Prager
682

Grout
100 to 2000
0.4
Solid ElemMohr Coulomb

Bolts
210000
0.3
8030
350
800
Solid ElemVon Mises

GEOMATE- Bangkok, Nov. 14-16, 2016

NORMAL CONSTRUCTION CASE
The first considered case is for normal
construction condition, where no accidental cases
occurred during construction. This model is used tocheck the value of deflection and to check any local
plastic points occurred in the model in ordinary cases.
Three load cases are considered in this category. In thefirst case, lining and grout layer are activated but themodules of elasticity of the grout is very small. Thissituation doesn’t allow direct contact between thelining and the soil. In the second case, only the load of
the grout and the own weight of lining and grout are
activated. The soil load is not applied because there is
still a gap between the lining and the soil. The gap isfilled with grout but in this case the grout is too week
to transfer the loads. In the third case the modulus ofelasticity of the grout is increased to its maximum
value in Table 1 which is the final value after
hardening. In the same time the connecting bolts are
removed from the ring. This case aims to study thepossibility of removing the connecting bolts after
hardening of the grout around the ring.

Fig. 5 Vertical deformation after hardening of grout
and removing bolts.

Analysis of Results
As shown in fig. (4-7), the deflection at crown oftunnel after construction of ring 2 is 21mm after ring
construction, where the bolts are installed. No plastic
zones generated around the segments except a very
local plastic zones generated around bolts. Due to thehigh tensile forces in bolts, bolts got deformed andplastic zones generated on the bolts. The settlement
slightly increases if bolts removed after hardening ofgrout, the increase is very small. Whereas, no plastic
concentration occurs after removing bolts. As aconclusion, after the hardening of grout there is noneed for bolts and can be safely removed from the ring.

Fig. 6 Distribution of plastic zones after hardening ofgrout and removing bolts.

Fig. 7 Plastic zones in the area of segment aroundconnecting bolt.
EFFECT OF BOLTS–A PARAMETRIC STUDY
During the erection of rings, sometimes TBM
machine crew does not erect some or all bolts between
the segments. This could happen by mistake orintentionally to reduce the consuming time or cost.
Erection of bolts has great effect on the stability ofrings during construction of rings until the hardeningof the surrounding grout layer. In this section, theeffect of not installing some or all bolts during
construction of the ring is investigated and examine towhich extent this case could lead to collapse of thering.

Fig. 4 Vertical deformation after construction of ring.

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The same load cases used in the previous section
are applied (The normal construction case). Numberof bolts in the ring are 8 bolts, one bolt for each joint
as shown in Fig. 8. In each analysis a number of bolts
are not activated.

Fig. 10 Distribution of plastic zones at bolt No. 8 for
not activating 1 bolt.
EFFECT OF EXCESSIVE GROUT ON RING
STABILITY
Grouting are used to fill the annular void
between the segmental lining and surrounding ground.
The interaction between machine, lining and the grout
is very important. Attention should be paid to avoid
excessive grout which causes destructive effect on
lining. Excessive grout could be generated from two
possible ways:
After grouting the lining, grouting volume will
settle, opining a gap on the upper part of cross section.
In this case more grouting is required to fill this gap,
this grouting called secondary grouting. Secondary
grouting is usually undertaken within 20m of the lastring built. It may cause destructive effect if its value
overestimated, which makes it concentrated by large
value in certain segments [11].
The other possible source of excessive grout
comes from the fact that grout pushed into the annularvoid between lining and ground through nozzle. If
some nozzles got clogged the grout around lining will
be redistributed and will be concentrated aroundcertain segments.
Two cases will be studied as follows:
1- Uniform excessive grout around the whole ring.
2- Excessive grout in some segments.
The study will consider increase in grout byvalue start from 10% of original value to 100% of theoriginal value.

Fig. 8 Numbering and orientation of bolts/joints.
Analysis Results

Deflection (cm)As shown in Fig. 9, longitudinal bolts have great
effect in the stability of the segmental lining during
construction. Even if one bolt only is not activated, thevalue of crown deflection increases by percentage of
39% of the normal case. Fig. 10 shows plastic zones at
Bolt No. 8 when only one bolt is not installed.

Number of not activated bolts

Fig. 9 Value of crown deformation versus the numberof not activated bolts.

Analysis of Results
As shown in Fig. 12, the main damage effect onring from excessive grout will happened in case ofsome segments only have excessive grout. As long as

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the excessive grout value increases the deflection will
increase until the failure case.
From the distribution of plastic zones shown in
Fig. (13-15), it is clear that in all accidental cases any
failure or plasticity zones generated first in bolts at thejoint between segments until failure. This leads to thefollowing suggestions:
1- Using nonlinear criteria for modeling concretesegmental lining have small importance, since most of
expected failure will occur at joints.
2- It is better to use two bolts between each twosegments to face any unexpected accidental cases, andreduce the risk of accidental cases on segments.
Effect of excessive grout in certain segments isvery destructive and one of major reasons which may
cause failure.The required grout pressure value should
be carefully calculated and grout injection process
should be carefully performed under careful
supervision to prevent any excessive pressure
generated on segments.

Deflection (cm)Fig. 14 Distribution of plastic zones at bolts foruniform excessive grout at all segments by 100%

Percentage increase of excessive grout(a)Fig. 12 Value of crown deformation versus the valueof excessive grout.

(b)
Fig. 15 Distribution of plastic zones for excessive
grout by 100% value at one segment:(a) At the edges of segments, (b) At bolt.

Fig. 13 Distribution of plastic zones at segments foruniform excessive grout at all segments by 100%685

GEOMATE- Bangkok, Nov. 14-16, 2016[2] Gruebl F., “Segmental ring design (Newchallenges with high tunnel diameters)”, World
Tunnel Congress 2012, Bangkok, Thailand,
Tunnelling and Underground Space for a Global
Society, 20 / 23 May 2012.
[3] Güray A., “Evaluation of structural analysis
methods used for the design of TBM segmental
linings”, Post Graduate Ms. Thesis, The
Graduate School of Natural and Applied
Sciences of Middle East Technical University,
Turkey, 2010.
[4] Hinchberger S.D., Ahn T., Lo K.Y.,“Investigation of segmental concrete tunnel liner
joint stiffness”, World Tunnel Congress and 36th
General Assembly, Vancouver, Canada, ITAAITES, 17 and 19 May, 2010.
[5] ITA WG2 on General Approaches to the Designof Tunnels, “Guidelines for the design of shieldtunnellining”,
International
Tunnelling
Association, Working Group No. 2, Research,
2000.
[6] ITA WG2 on General Approaches to the Designof Tunnels, “Guidelines for the design of tunnels
secondreport”, International Tunnelling
Association, Working Group No. 2, Research,
1988.
[7] JSCE, “Japanese standard for shield tunnelling,
Japan Society of Civil Engineers”, The fourth
edition, Tokyo, 2000, English translation 2007.
[8] Mahajan R., “3 Dimensional Structural
Modelling of Segmental Tunnel Lining”,
Australian Tunnelling Society (ATS) Journal,
October 2010.
[9] MIDAS, MIDAS GTS NX V3.0.0 (4), MIDASGTS Manual, MIDAS Information Technology
Co., Ltd, Gyeonggido, Korea.
[10] NAT, National Authority for Tunnels. Greater
Cairo Metro Line III, Project Documents, Cairo,
Egypt, 2009.
[11] Toan N. D., “TBM and lining, essential
interfaces”, Post Graduate Ms. Thesis,
Politecnico Di Torino, 2006.
[12] U.S Department of transportation – Federal
Highway administration (U.S.A), “Technicalmanual for design and construction of roadtunnels”, Civil Elements, Publication No.
FHWA-NHI-, December 2009, Pp. 10-34.

CONCLUSIONS
Several cases have been investigated to study the
behavior of TBM segmental lining at accidental cases
during construction. From these cases the following
conclusions may be drawn: After ring construction, where the bolts installed,
no plastic zones generated around the segments except
very local plastic zones generated around bolts.
 In normal cases and due to the tensile forces inbolts, bolts got deformed and plastic zones generated
on them.
 The settlement slightly increases if bolts removed
after hardening of grout, the increase is very small,whereas, no plastic concentration occurs after
removing bolts.
 After the hardening of grout there is no need forbolts and it can be safely removed from ring.
 Installation of bolts is very important during theconstruction of lining and should be carefully handled.
 If more than one bolt isn’t installed during
construction of the ring, this could lead to ring failure.
 Excessive grout concentrated at some rings has
destructive effect on the ring. This situation could leadto collapse of ring. Consequently this situation should
be avoided during construction of TBM segmental
lining.
 Uniform excessive grout around the whole ring
have small effect on the stability of ring. Especially incases where the value of excessive grout is small.
Noting that, the case of uniform excessive grout israrely happened and normally the excessive grout
concentrated in certain segments.
 From the distribution of plastic zones shown
earlier, it is notable that in all accidental cases, any
failure or plasticity zones generated first in bolts at thejoint between segments. Consequently it is better to
use two bolts between each successive segments tominimize any unexpected accidental cases.
REFERENCES[1] Duddeck H., Erdmann J., “Structural designmodels for tunnels”, Scribd wbsite 2012,
Universitat Braunschweig, Germany, 1982.

686

Sixth International Conference on Geotechnique, Construction Materials andEnvironment, Bangkok, Thailand, Nov. 14-16, 2016, ISBN: 978-4-9905958-6-9 C3051

EFFECT OF THICKNESS ACOUSTIC PANELS UTILIZING COCONUT
COIR
D.Salinah1, M.T.Fadzlita1 , G.Habibah1 , L.Janice Ayog1, A.Adriana1 and A.Hassanel1
1
Faculty of Engineering, University Malaysia Sabah, Malaysia

ABSTRACTThis research was carried out to study the effect of thickness acoustic panels of coconut coir to its acoustic
properties. The acoustic panels were tested for its acoustical properties to determine its acoustical performanceand characteristics. The coconut coir was tested with thickness of 10 mm, 20 mm, 30 mm and 40 mm with aconstant area of 6 m2. Two types of test were conducted toward the specimens, which are, the stimulation test
using the WinFLAGTM and the reverberation room experimental test. The stimulation is used to predict the resultof the sound absorption coefficient in the reverberation room. All specimens were tested at the acoustic
laboratory of Faculty of Engineering University Malaysia Sabah accordance to ISO 354 (1985) standard fornoise absorption coefficients at frequencies of 125 Hz, 250 Hz, 500 Hz, 1000 Hz, 2000 Hz and 4000 Hz. The
result shows that coconut coir has better sound absorbing quality than fiberglass for low frequency at 500 Hz andbelow. Hence, coconut coir has high potential to be applied as sound absorbing material.
Keywords: Acoustic; Thickness; Coconut coir; Sound absorption coefficient.

which is, they all have porous surface. This
characteristic, enable them to be used in acousticalappliances such as acoustic panels where they could
absorb and insulate sound that propagates at them[9].
Previous researches discovered that coconut coirfibers have a good potential as sound absorbing
panel as it have a good sound absorbing properties
for a wide frequency range [10]-[12].

INTRODUCTION
Sound absorption panels are panels design andused for the purpose of absorbing sound as noise
control. Generally, noise control is critical aspect inany structural and civil engineering designs.
Acoustical appliances which in this case soundabsorbing acoustic panel, synthetic fiber-basedmaterial such as rock wool and fiberglass are widelyused as sound absorbent in acoustic panels. The
problem with these acoustic panel material are its
health hazard issues if exposed in long period. Forfiberglass and other synthetic-fiber based materials,
there are growing concerns related to health andsafety issues due to the potential health hazards thesematerials possess. The shredding of these materials
can be detrimental to the health when exposed to theeyes or lungs for instance when their fibers areinhaled, since they can lay down in the lung alveoli.
In addition, they can cause skin irritation [1].

Many previous researches have been made in thestudies of the effect of material thickness to thequality of absorption of the material. Everest [13]states that, for the lower frequencies it is logical toexpect greater sound absorption from thicker
materials. Meanwhile, higher frequency has inverse
effect of declining absorption coefficient withincreasing thickness of materials.
Thus, it is a fact that this research will bring
benefit to various people. It is a necessity to fully
utilize the natural fibers to replace the used ofsynthetic fiber glass that are non- environmental andhealth friendly. Furthermore, there is a lack of acomprehensive, validate and easily accessible
database on effect of thickness on coconut coir fiber
panel
to
its
acoustic
performanceandcharactericticsas related to its structural applications.
Therefore, the aim of this research is to study thepotential use of coconut coir fiber as soundabsorption panel and the effect of thickness on
sound absorption coefficient of acoustic panel was
also studied.

This give opportunity for alternative materials
from organic fibers to be developed as a replacement
materials for producing sound absorbing panels.
Thus, many research have been carried out todiscovered potential usage of new materials forsound absorption panel. Many researches focused on
use of natural fiber due to its advantages in sense ofcost, abundant and no risk in health and safety [2][8].
Natural fibers such as oil palm fiber, coconut coir,wood husk and paddy husk have a common
characteristic that is crucial for acoustical appliances
687

GEOMATE- Bangkok, Nov. 14-16, 2016

EXPERIMENTAL DETAILS
Materials
Two types of sample were tested to determine
their sound absorption coefficient which is glass
fiber and coconut coir fiber. Readily made glass
fiber of 1 m2 size each, obtained from Acoustic
Laboratory, Faculty of Engineering. Coconut coirfiber of 1 m2 size each is obtained from coconut coirmanufacturing company with different thickness.
The coconut coir fiber mat sample was ordered froma local coconut coir manufacturer that produces thefiber for furniture appliances. The coconut coir matsdimension is approximately 1 m in length and 1 m inwidth. In this research, the effect of thickness to thesound absorption coefficient were studied. The
coconut coir was tested with thickness of 10 mm, 20mm, 30 mm and 40 mm with a constant area of 6 m2.
The total weight of the specimen is approximately
10.5 kg and the density is approximately 108.03
kg/m3. In this research, the type of fiberglass used is
Woven fiberglass Type T, which is used forcommercialized thermal and sound insulationmaterial. The fiberglass was tested to determine its
sound absorption coefficient which then comparedwith coconut coir fiber.

Fig. 2 Woven Fiberglass mat Type T.
Test Conditions
Test was conducted by reverberation room
method and simulation using WinFLAGTM. In thisresearch, coconut coir fiber will be used as soundabsorbing material. The area of the test specimen isapproximately 6m2 which covers 20% of the total
area of the reverberation room floor. The areas of thetest specimen were constant for every test. The only
parameters that were manipulated are the thicknessof the test specimen. Series of random noise withdifferent frequencies were used in the test. The
absorption of the reverberation room is measured as
outlined in both before and after placing a specimenof material to be tested in the room. The increase inabsorption is divided by the area of the test specimen
to obtain the dimensionless sound absorption
coefficient. Sound absorption coefficients for thepanels to be tested were determined by reverberation
room method accordance to ASTM C423-02a(Standard Test Method for Sound Absorption
Coefficients by the Reverberation Room Method)and ISO 354. The overall internal working
dimensions of the room were 7150 mm length, 4100
mm width, and 5150 mm height. The working
volume of the room, V is approximately 151 m3.
Measurement, analyzing and calculation ofreverberation time and value of sound absorption
coefficient were done using dBBATI32 softwaredeveloped by dB STELL. In this research,
WinFLAGTM was used to simulate and calculates
sound absorption coefficient of porous absorber. The
result of this simulation was then compared to theresult obtained from reverberation room method.

Fig. 1 Coconut coir test specimen.

688

GEOMATE- Bangkok, Nov. 14-16, 2016

RESULT AND DISCUSSION
The result consisted of stimulation data andexperimental data. The stimulation data wereobtained through the use of WinFLAGTM. Table 1
and 2 show the stimulation (S) and experimental (E)test result obtained for coconut coir panel andfiberglass.
Comparison of Sound Absorption Characteristic
between Fiberglass and Coconut Coir Fiber
Table 1 Stimulation and experimental results for
Fiberglass and coconut coir fiber panel sample

125
250
500
1000
2000
4000

Sound Absorption Coefficient, αFiberglass
Stimulation
Experimental
0.04
0.06
0.19
0.11
0.51
0.56
0.68
0.82
0.75
0.77
0.84
0.61

125
250
500
1000
2000
4000

Coconut Coir Fiber
Stimulation
Experimental
0.08
0.10
0.26
0.15
0.53
0.42
0.67
0.72
0.71
0.66
0.78
0.51

f (Hz)
Fig. 3 Schematic and dimension of the reverberation
room without test specimen.

Fig. 4 Reverberation room with 20 mm thickness
6 m2 area of coconut coir fiber.

Table 2 Stimulation and experimental results forcoconut coir fiber panel sample - Different
Thicknessf (Hz)

125
250
500
1000
2000
4000
Fig. 5 Reverberation room with 20 mm thickness
6 m2 area of fiberglass

125
250
500
1000
2000
4000

689

Sound Absorption Coefficient, α
10 mm
20 mm
S
E
S
E
0.04
0.05
0.06
0.10
0.12
0.10
0.26
0.15
0.28
0.34
0.53
0.42
0.44
0.65
0.67
0.72
0.53
0.68
0.71
0.66
0.62
0.52
0.78
0.51
30 mm
40 mm
S
E
S
E
0.13
0.16
0.19
0.27
0.41
0.28
0.57
0.41
0.73
0.58
0.87
0.69
0.78
0.75
0.85
0.81
0.80
0.63
0.86
0.62
0.85
0.49
0.90
0.51

GEOMATE- Bangkok, Nov. 14-16, 2016

Table 1 shows that at low frequency range from0 to 500Hz, sound absorption coefficient value forboth material increases linearly. The patternindicates that for low frequency of 125 Hz, 250 Hz,
and 500 Hz, coconut coir fiber have better soundabsorption coefficient than fiberglass. For higher
frequency of 1000 Hz, sound absorption coefficient,
α for fiberglass increase to 0.68 which is higher thancoconut coir fiber with sound absorption coefficient,
α of 0.67. Fiberglass reaches it maximum soundabsorption coefficient, α at 4000 Hz with value of0.84 and coconut coir fiber with maximum of 0.78 at
4000 Hz.

quality for every frequency band
Based on table 2, all four test specimens havealmost the similar pattern where sound absorption
coefficient, α for all test specimens increases withincreasing frequency. Maximum sound absorption
coefficient, α for coconut coir with 10 mm thickness
is at 4000 Hz with value of 0.62. Coconut coir withother thickness also reaches its maximum soundabsorption coefficient, α at 4000 Hz with value of0.78 for 20 mm, 0.85 for 30 mm and 0.9 for 40 mm.
Also, at all frequency coconut coir with larger
thickness have larger sound absorption coefficient, α.
At low frequency, 125 Hz, all four test specimens
have very low sound absorption coefficient, α withvalue below 0.2. Hence from this simulation it canbe concluded that larger thickness gives larger soundabsorption coefficient, α thus higher soundabsorption quality.

Clearly that for low frequency of 250 Hz andbelow, sound absorption coefficient, α for coconut
coir fiber is higher than fiberglass, but for higher
frequency of 500 Hz until 4000 Hz, fiberglass havehigher sound absorption coefficient, α than coconut
coir fiber. From the relation of λ = C/ƒ where ƒ isfrequency, λ is wavelength, and C is speed of soundwhich is 3.00 x 108 m/s it can be deduce that low
frequency have longer wavelength. With longer
wavelength, the sound wave can easily travels inside
the larger pores of coconut coir fiber rather thanfiberglass that have smaller pores. The sound waveof low frequency can travel into coconut coir fiberand vibrates with the fibers surface more easily tocompare with fiberglass thus converting the soundenergy into heat energy. As results, it gives coconut
coir fiber a higher sound absorption coefficient, α forlow frequency [13].

It can be observed that sound absorption
coefficient, α increase from 125 Hz and keepincreasing until maximum point at 1000 Hz for 20
mm, 30 mm, and 40 mm but 2000 Hz for 10 mm.
For frequency above 1000 Hz, sound absorption
coefficient, α for all four test specimens started todecrease to average value of 0.5. For 10 mm coconut
coir thickness, minimum sound absorption
coefficient, α is 0.05 and maximum soundabsorption coefficient, α is 0.68. For 20 mm coconut
coir thickness, minimum sound absorption
coefficient, α is 0.1 and maximum sound absorption
coefficient, α is 0.72 which is higher than 10 mm
coconut coir thickness. For coconut coir with 30 mm
thickness, minimum sound absorption coefficient, α
is 0.16 and maximum sound absorption coefficient,
α is 0.75, higher than 10 mm and 20 mm coconut
coir thickness. For coconut coir with 40 mm
thickness, minimum sound absorption coefficient, α
is 0.27 and maximum sound absorption coefficient,
α is 0.81 which is the highest value to comparedwith other thickness.

For higher frequencies, it has smaller wavelength
according previous relation. With much more
smaller wavelength, the sound wave can travels into
small fiberglass pores and convert sound energy toheat energy thru vibration more effectively. Coconut
coir fiber has larger pores, thus it unable to absorb
effectively sound wave with smaller wavelength
because it has more gap in interstices of fiber
making collision of sound wave with coconut coirsurface much less to compare with much tightly
packed structure of fiberglass [14].

For 10 mm coconut coir thickness, soundabsorption coefficient, α at 2000 Hz and 4000 Hz
decrease to 0.68 and 0.52 respectively. Meanwhile,for 20 mm thickness, sound absorption coefficient, α
at 2000 Hz and 4000 Hz also decrease to 0.66 and0.51 respectively. For 30 mm coconut coir thickness,sound absorption coefficient, α also decrease at same
point to 0.63 and 0.49. Meanwhile, for 40 mm
thickness, sound absorption coefficient, α at 2000 Hz
and 4000 Hz decrease to 0.62 and 0.51 respectively.

Effect Of Coconut Coir Thickness To Sound
Absorption Characteristics Of Coconut Coir
Acoustic Panel
The effect of coconut coir fiber thickness wasstudied to observed the behavior of sound absorption
characteristic of coconut coir when different
thickness is use. Test is conduct by reverberation
room method and simulation using WinFLAGTM.
Four tests were conducted with coconut coir having
different thickness of 10 mm, 20 mm, 30 mm, and
40 mm. All four test specimens have constant area of
6 m2. This test was also done to studied the optimum
thickness that gives the best sound absorption

This shows that for low frequency band ranging
from 100 Hz to 1000 Hz, thicker coconut coir gives
higher sound absorption coefficient, α means, bettersound absorption quality. For higher frequency
above 1000 Hz, sound absorption coefficient, α690

GEOMATE- Bangkok, Nov. 14-16, 2016

decrease with increasing thickness. Hence, for highfrequency application, thicker coconut coir is notrecommended. This theory is supported by previous
research by Everest [13] stated that it is logical toexpect greater sound absorption from thicker
materials, but this logic holds primarily for the lower
frequencies.

Shaji, J. and Thomas, S. Coir and Hemp
FiberReinforced Polymer Composites. In
Fakirov, S. and Bhattacharyya D. (Eds.) 2007.
Handbooks of Engineering Biopolymers:
Homopolymers, Blends and Composites, pp.
285-309. Munich: Hanser Gardner Publications,Inc.
[5] Joshi, S. V., Drzal L. T., Mohanty, A. K. andArora, S. 2003. Are natural fiber composites
environmentally superior to glass fiber
reinforced composites?. Composites Part A:
Applied Science and Manufacturing 35(3): pp.
371-376.
[6] Satyanarayana, K. G., Kulkarni, A. G. andRohatgi, P. K. 1981. Structure and properties ofcoir fibres. Sadhana 4(4): pp.419-436.
[7] Sivapragasam, A. 2008. Coconuts in Malaysia –current developments and potential for
revitalization. Shah Alam: MARDI[8] Wambua, P., Ivens, J. and Verpoest, I. 2003.
Natural fibres: can they replace glass in fibre
reinforced plastics?. Composite Science andTechnology 63(9): pp.1259-1264.
[9] Clemons, Craig M., Caulfield, Daniel F. 2005.
Natural Fibers: Functional Fillers for Plastics.
Weinheim: Wiley-VCH, 2005: pp.195-206.
[10]Fouladi, M. H., Ayub, Md. and Nor, M. J. M.
2011. Analysis of Coir Fiber Acoustical
Characteristics. Applied Acoustics 72 (1):pp.35-42.
[11]Mahzan, S., Zaidi, A. M. A., Arsat, N., Hatta, M.
N. M., Ghazali, M. I. and Mohideen, S. R. 2010.
Study on sound absorption properties of coconut
coir fiber reinforced composite with added
recycle rubber. International Journal of
Integrated Engineering (Issue on Mechanical,
Materials and Manufarcturing Engineering 2(1):pp.29-34.
[12]Ayub, Md., Nor, M. J. M., Amin. N., Zulkifli. R.
and Ismail, A. R. 2009. A Preliminary Study of
Effect of Air Gap on Sound Absorption of
Natural Coir Fiber, Proceeding of the Regional
Engineering Postgraduate 2009.
[13]Everest,F,A. 2001. Master Handbook of
Acoustic. New York: McGraw-Hill.
[14]Bucur,V. 2006. Acoustic of Wood. Berlin:
Springer Publications.

CONCLUSION
Experimental result shows that coconut coir has
better sound absorbing quality than fiberglass for
low frequency at 500 Hz and below. Therefore,
coconut coir has high potential to be applied assound absorbing material. By using coconut coir as
substituting the use of other commercial porous
sound absorbing material, cost can be reduce as
coconut coir fiber is easier to obtain, cheaper andeasier to process. Moreover, coconut coir fiber issafe and not hazardous to humans compare tofiberglass or other synthetic fibers. Meanwhile,sound absorption coefficient increases withincreasing of thickness. However, it holds only for
low frequency range. The effect of test specimen
area to sound absorption characteristics of coconut
coir acoustic panel is also studied in this research.
Result shows that lower area of test specimen gives
better sound absorption. This is due to the diffraction
effect of sound energy around the perimeters of thesound absorbing materials and to the additional
absorption provided by the exposed panel edges. It
can be concluded that coconut coir show highpotential to be applied as sound absorbing material.
This natural fiber can be used to replace synthetic
fiber and mineral wool in the context of acoustics
environment and control.
ACKNOWLEDGEMENTS
The author of this paper would like to express
their gratitude and sincere appreciation to University
Malaysia Sabah for providing the financial support.
REFERENCES[1] Asdrubali, F. 2006. Survey on the Acoustical
Properties of New Sustainable Materials for
Noise Control. Proceedings of Euronoise 2006.
30 May - 1 June, 2006. Tampere, Finland,. pp.
1- 10.
[2] Yang, H. S., Kim, D. J. and Kim, H. J. 2003.
Rice straw-wood particle composite for soundabsorbing wooden construction materials.
Bioresource Technology 86(2): pp.117-121.
[3] Hong, Z., Bo, L. Guangsu, H. and Jia, H. 2007.
A novel composite sound absorber withrecycled rubber particles. Journal of Sound andVibration 304(1): pp.400-406.
[4] John, M. J., Indira, K. N., Geethamma, V. G.,
691

Authors Index

A

B

A. W. Hago

455

Bambang Suhendro

210

Abdul Halim Abdul L.

635, 557

Ben W. Kolosz

265

Bernardo A. Lejano402, 408, 449

Abdul Karim M. Zein
Abdul Naser Abdul G.

245

Brian Lamb

22

Adnan Zainorabidin

326

Bryan Josef T. Medrano

646

Adnène KASSEBI

538

Buddhima Indraratna

156, 34

Agawit Thaothip

257

Agustinus Isjudarto

660

Ahmad Hilmy Abdul H.

245

C

Ahmad Rifa’i

210

Cesar Palma Jr. III

390

Akiho Inoue

52

Chanchai Submaneewong

40

Akira Kondo

578

Chayanon Hansapinyo

512

Akira Tomioka

46

Cherdsak S.

67１

415

Chinapat Buachart

512

Alexander Goudov

665

Cholachat Rujikiatkamjorn

156, 34

Ali Nisari Tabrizi

286

Choy Soon Tan

270

Ali Sobhanmanesh

270, 675

Chris Walker

494

Alina Paranina

377

Christophe Balg

373

Amin Esmail Khalil

557

Clevin John V. Santos

186

Aminaton Marto

270

Cong-Oanh Nguyen

308

Anant Aishwarya Dubey

103

Anas Puri

210

D

Anusron Chuesamat

355

Daisuke Hayasaka

532

Arief Rachmansyah

64Danai Thaitakoo

618

Artit Udomchai

367, 675

David A.C. Manning

265

Arul Arulrajah

2, 396, 438, 671

Davin H. E. Setiamarga

517

Atsushi Iizuka

233

Dennis Sindico390

Atsushi Yashima

46, 52

Deprizon Syamsunur

675

Avirut Chinkulkijniwat

227, 367

Dilum Dissanayake

265

Duc Bui Van

227, 367

Aldwin Christopher

A. L.

Authors Index

E

Hirosuke Hirano

517

Erica Elice Saloma Uy

426, 432

Hirotaka Saito

355

Euclid de Guzman390

Hiroto Akimoto

500

Hiroyuki Araki

320

Hiroyuki Ii

618

Hita Pandita

660

F
Farshid Maghool

438

Feng Zhang

151

Fouad Zargouni

538

I

Fumio Tatsuoka

443

Indra Hardi

331

Fumitake Nishimura

610

Irin Limrat

367, 675

Isamu Yoshitake

420

Ismael Concepcion Jr.

168

Itthikorn Phummiphan

2

G
Georgios Katsigiannis

58

Gilford B. Estores

408

Gyanendra Sthapit

455

H

J
J. N. Mandal

485

Jana Labudkova

473

H. Lahuta

343

Jay Lobwein

22

Habib Musa Mohamad

326

Jayvee L. Gagan

402

Hary C. Hardiyatmo

210

Jecelle Elyssa E. Calip

415

Hasan M. Faisal

349

Jeerapan Donrak

2

Hassanel Zachary A.

337

Jemison L. Go

468

Hayat Bouchum

461

Jim Shiau

22

Hayato Ishigaki

320

Joenel G. Galupino

506

Hermie M. del Pilar

186, 415

John L. Tan

468

Hidenori Tanaka

384

Jonathan R. Dungca

28,168, 426, 506

Hideo Noguchi

629

Joshua A. M. Collado

468

Hideyuki Ito

629

Julie Ann L. Jao

28

Hikari Shimadera

578

Julinda Hena J.

337

Hiroaki Hasegawa

420

Jun Tanaka

532

Hiroshi Tsuno

545

Jun-ichi Yamazaki

135Authors Index

K

M. Mohyla

343

K. Ruengwirojjanakul

92

M. Pinka

343

K. Vojtasik

343

M. Stolarik

343

K. Watcharasawe

71

M. Sugimoto

98

Karin Kandananond

640

Mahad Baawain

455

Katsuhiko Koizumi

320

Mai Chinzaka

578

Katsuyuki Kawai

233

Makoto Yamaguchi

205

Kayo Doumoto

205

Marion Ryan Vicencio

168

Kazuki Mihara

600

Marisol E. Manarin

186

Kazuki Nosho

129

Mark A. Goddard

265

Kazunari Tanaka

652, 656

Mary Ann Q. Adajar

390, 432

Kazunori Nakashima

145

Masaaki Katagiri

205

Kazuya Itoho

500

Masaki Kitazume

192

Keeratikan Piriyakul

276

Masanobu Taniguchi

618

Ken-ichi Shishido

123

Masato Nakamichi

205

Kentaro Kondo

532

Mathew Sams

22

Kentaro Uemura

77, 162

Md Yeasin Ahmed

563

Khalid Ahmad

245

Meng Jing

239

Khalifa S. Al-Jabri

455

Menglim Hoy

2

Kiyonobu Kasama

139, 205

Michiko Masuda

610

Koichi Nagao

162

Mikako Ishii

569

Koichi Yamanaka

629

Mike A. Rosanto

468

Kouki Zen

205

Minoru Yamanaka

320

Krit Won-In

479

Misa Konishi

604

Krizza Diane M. Santos

186

Misato Sekitani

123

Kunio Minegishi

629

Mitsunari Hirasawa

139, 205

Kyoichi Okamoto

517, 584, 590

Moeko Matoba

10

Mohamed Ayeldeen

192

Mohamed Tahiri

461

Mohammad Reza Atrchian286

Mohd Yuhyi Mohd Tadza

215

Moises Christian Mickail L.

168,

L
Lessandro Estelito O. G.

646

M
M. Ehsan Jorat

265

Monzur Imteaz563

M. Faisal

623

Muhd Shahril Nizam Ismail

245

Authors Index

N

R

N. Heama

92, 117

Ramli Nazir

675

Naoaki Suemasa

162, 174, 500

R. Nuntasarn

281

Neil Patrick C. Nase

415

R. Sukkarak

198

Ngoc Trung Ngo

34

Radim Cajka

473

Nishant Neeraj

293

Rafiqul A. Tarefder349

Nobuhiro Hisabe

420

Rajeshwar Goodary

227

Nobuyuki Nishimiya

517

Ramanathan A.

293

Nobuyuki Soga

46, 52Raul Fuentes

58

Nuchnapa Tangboriboon

551

Richard M. De Jesus

468

Nurhidayah Mahazam

215

Robert Evans

563

Nurten Akgun

265

Rusutsu Ito596

Nutapong Hirano

314

Rutchakit Maythathirut

671

Nutthapon Thasnanipan

361

Ryan Ho390

Ryo Hashimoto

135

O
Osamu Kawahara

205

P

S
S. Suwansawatt92, 117

Safawati Mohd Radzi

245

P. Chaipanna

98

Sahapap Supawo

361

P. Jongpradist

71, 92, 98, 117, 198

Saori Iwamoto

569

P. Kitiyodom

71

Saran P. Sohi

265

P. Lueprasert

92, 117

Satoru Kawasaki

145

Paresh575

Sawanya Dararat

250

Pedro Ferreira

58

Seiji Kobayashi

151

Peter Schwammberger

494

Sereyroath Chea

361

Phan Do Hung

545

Sermsak Tiyusangthong

671

Pisut Rodvinij

303

Shin Yoshikawa

652, 656

Pitiwat Wattanachai

303

Shinya Inazumi

123, 129, 135

Pornkasem Jongpradist

40

Shiori Yanagawa

604

Prasanna P.Kulkarni

485

Shota Inoue

139

Putli Yasmeen N. M. Y.

337

Shugo Morita

517

Shuichi Hasegawa

320

Vasantlal Dalal

Authors Index

Shumpei Mitsuyama

145

Tanwee Mazumder

293

Shunsuke Shimada

162

Tatsuhide H.

545

Shuuichi Kuwahara

129

Tawatchai T.

314

Sirikanya Laosuwan

314

Tenshiro Goto

10

Siriwan Waichita

40

Terence Oliver See

168

Siti Aishah M. S.

337

Terry Lucke

494

Snehasis Tripathy

215

Teruo Arase

522, 526, 532

Somjai Yubonchit

227,367,675

Tetsuo Okano

522, 526, 532

Souad Zyade

461

Tetsuoh Shirota

522, 526

Sukartono

660

Tetsuya Yanobe

652

Sukhveender S. S. S.

635

Thanawat Meesak

479

Suksun Horpibulsuk

2, 367,396, 438, 671

Thanh Thi Tran

308

Suresh Kumar

103

Thanh Trung N.

156

Suriyon Prempramote

83

Thayanan B.

361

Suttisak Soralump

123, 129, 135

Thitapan C.

443

Suvimol Sujjavanich

479

Thomas H.

373

Tin Nguyen

22

Tomoe Komoriya

584, 590

T
TA. Kua

396

Tomonori K.

604

Tadamasa Fukagawa

610

Tomoyuki Mizuta

135

Taizo Uchida532

Toshihiko Matsuto

569

Takahiro Fujii

629

Toshikazu Hori

355

Takamitsu Sasaki

162, 174

Toshiro Hata

596, 600

Takashi Hara46, 52

Toshiyuki Nagano

205

Takashi Shinsaka

135

Tsutomu N.

180

Takashi Toyama

590

Tsuyoshi Tanaka

500

Takashi Tsunemoto

151

Tulasi Ram B.

367

Takashi Umeyama

205

Takayuki Matsuo

569

Takeshi Toyama

517, 584

U

Takumi Sakai

656

Umme A. Mannan

349

Takuya Kawakatsu

233

Usmanov R. A.

299

Tan Choy Soon

675

Authors Index

V

Yoshihiro N.

135

VShunsuke Takiura

174

Yoshihiro Yokota

52

Valeriy Perminov

665

Youssef Halimi

461

Van-Tram Thi Dao

308

Yuhei Kurimoto

151

Viggo Jensen

479

Yuichi Yahiro

205

Yuji Hara

618

Yuji Kohgo

355

W
W. Wannakul

281

Yuki Hara

192

Waleed Abdul K.

221

Yuki Imai

604

Wanitcha Unjan

551

Yuki Tomoguchi

532

Warat Kongkitkul

250, 257, 443

Yukoh Hasunuma

77, 162

Watcharagon W.

479

Yulvi Zaika

64

Werasak Raongjant

239

Yusuke Kuroda

420

Yusuke Torigoe

129

Yuto Uragaki

517

Yuuya Nakase

135

Y
YJ. Du396, 438

Yasuhide Mochida

331

Yasumasa Tojo

569

Z

Yoshifumi Taguchi

111

Zaw Zaw Aye

361

Yoshihiro Kimura

10

Zentaro Furukawa

205

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