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Chapter 9 Commentary 
CONCRETE STRUCTURE DESIGN REQUIREMENTS 
9.1 GENERAL 
9.1.2 References. The main concern of Chapter 9 is the proper detailing of reinforced concrete 
construction for earthquake resistance. The bulk of the detailing requirements in this chapter are 
contained in ACI 318. The commentary for ACI 318 contains a valuable discussion of the rationale 
behind detailing requirements and is not repeated here. 
9.2 GENERAL DESIGN REQUIREMENTS 
9.2.1 Classification of shear walls. In the 2000 Provisions, shearwalls are classified by the amount and 
type of detailing required. This classification was developed to facilitate assigning shearwalls to seismic 
design categories. 
9.2.2 Modifications to ACI 318. The modifications noted for ACI 318 are: changes in load factors 
necessary to coordinate with the equivalent yield basis of this document, additional definitions and 
provisions necessary for seismic design requirements for structural systems composed of precast 
elements, and changes that incorporate certain features of the detailing requirements for reinforced 
concrete that have been adopted into the 1997 Uniform Building Code and the 2000 International 
Building Code. 
Procedures for design of a seismic-force-resisting structural system composed of precast elements 
interconnected predominately by dry joints require prior acceptance testing of modules of the generic 
structural system because with the existing state-of-knowledge, it is inappropriate to propose code 
provisions without such verification. 
The complexity of structural systems, configurations, and details possible with precast concrete elements 
requires: 
1. Selecting functional and compatible details for connections and members that are reliable and can be 
built with acceptable tolerances; 
2. Verifying experimentally the inelastic force-deformation relationships for welded, bolted, or grouted 
connections proposed for the seismic resisting elements of the building; and 
3. Analyzing the building using those connection relationships and the inelastic reversed cyclic loading 
effects imposed by the anticipated earthquake ground motions. 
Research conducted to date (Cheok and Lew, 1991; Elliott et al, 1992; Englekirk, 1987; French et al, 
1989; BSSC, 1987; Hawkins and Englekirk, 1987; Jayashanker and French, 1988; Mast, 1992; Nakaki 
and Englekirk, 1991; Neille, 1977; New Zealand Society, 1991; Pekau and Hum, 1991; Powell et al, 
1993; Priestley, 1991; Priestley and Tao, 1992; Stanton et al, 1986; Stanton et al, 1991) documents 
concepts for design using dry connections and the behavior of structural systems and subassemblages 
composed of precast elements both at and beyond peak strength levels for nonlinear reversed cyclic 
loadings. 
Use of prestressing tendons. Sec. 9.2.2.1.4 defines conditions under which prestressing tendons can be 
used, in conjunction with deformed reinforcing bars, in frames resisting earthquake forces. As 
documented in Ishizuka and Hawkins (1987), if those conditions are met no modification is necessary to 
the R and Cd factors of Table 4.3-1 when prestressing is used. Satisfactory seismic performance can be 
obtained when prestressing amounts greater than those permitted by Sec. 9.2.2.1.4 are used. However, as 
documented by Park and Thompson (1977) and Thompson and Park (1980) and as required by the 
combination of New Zealand Standards 3101:1982 and 4203:1992, ensuring satisfactory performance 
requires modification of the R and Cd factors. 
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9.3 SEISMIC DESIGN CATEGORY B 
Special details for ductility and toughness are not required in Seismic Design Category B. 
9.3.1 Ordinary moment frames. Since ordinary frames are permitted only in Seismic Design 
Categories A and B, they are not required to meet any particular seismic requirements. Attention should 
be paid to the often overlooked requirement for joint reinforcement in Sec.11.11.2 of ACI 318. 
9.4 SEISMIC DESIGN CATEGORY C 
A frame used as part of the seismic-force-resisting system in Seismic Design Category C is required to 
have certain details that are intended to help sustain integrity of the frame when subjected to deformation 
reversals into the nonlinear range of response. Such frames must have attributes of intermediate moment 
frames. Structural (shear) walls of buildings in Seismic Design Category C are to be designed in 
accordance with the requirements of ACI 318. 
9.4.1.1 Moment frames. The concept of moment frames for various levels of hazard zones and of 
performance is changed somewhat from the provisions of ACI 318. Two sets of moment frame detailing 
requirements are defined in ACI 318, one for regions of high seismic risk and the other for regions of 
moderate seismic risk. For the purposes of this document, the regions are made equivalent to Seismic 
Design Categories in which high risk means Seismic Design Categories D and E and moderate risk 
means Seismic Design Category C. This document labels these two frames the special moment frame 
and the intermediate moment frame, respectively. 
The level of inelastic energy absorption of the two frames is not the same. The Provisions introduce the 
concept that the R factors for these two frames should not be the same. The preliminary version of the 
Provisions (ATC 3-06) assigned the R for ordinary frames to what is now called the intermediate frame. 
In spite of the fact that the R factor for the intermediate frame is less than the R factor for the special 
frame, use of the intermediate frame is not permitted in the higher Seismic Design Categories (D, E, and 
F). On the other hand, this arrangement of the Provisions encourages consideration of the more stringent 
detailing practices for the special frame in Seismic Design Category C because the reward for use of the 
higher R factor can be weighed against the higher cost of the detailing requirements. The Provisions also 
introduce the concept that an intermediate frame may be part of a dual system in Seismic Design 
Category C. 
The differences in the performance basis of the requirements for the two types of frames might be 
summarized briefly as follows (see the commentary of ACI 318 for a more detailed discussion of the 
requirement for the special frame): 
1. The shear strength of beams and columns must not be less than that required when the member has 
yielded at each end in flexure. For the special frame, strain hardening and other factors are 
considered by raising the effective tensile strength of the bars to 125 percent of specified yield. For 
the intermediate frame, an escape clause is provided in that the calculated shear using double the 
prescribed seismic force may be substituted. Both types require the same minimum amount and 
maximum spacing of transverse reinforcement throughout the member. 
2. The shear strength of joints is limited and special provisions for anchoring bars in joints exist for 
special moment frames but not intermediate frames. Both frames require transverse reinforcement in 
joints although less is required for the intermediate frame. 
3. Closely spaced transverse reinforcement is required in regions of potential hinging (typically the ends 
of beams and columns) to control lateral buckling of longitudinal bars after the cover has spalled. 
The spacing limit is slightly more stringent for columns in the special frame. 
4. The amount of transverse reinforcement in regions of hinging for special frames is empirically tied to 
the concept of providing enough confinement of the concrete core to preserve a ductile response. 
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These amounts are not required in the intermediate frame and, in fact, for beams stirrups may be used 
in lieu of hoops. 
5. The special frame must follow the strong column/weak beam rule. Although this is not required for 
the intermediate frame, it is highly recommended for multistory construction. 
6. The maximum and minimum amounts of reinforcement are limited to prevent rebar congestion and to 
assure a nonbrittle flexural response. Although the precise limits are different for the two types of 
frames, a great portion of practical, buildable designs will satisfy both. 
7. Minimum amounts of continuous reinforcement to account for moment reversals are required by 
placing lower limits on the flexural strength at any cross section. Requirements for the two types of 
frames are similar. 
8. Locations for splices of reinforcement are more tightly controlled for the special frame. 
9. In addition, the special frame must satisfy numerous other requirements beyond the intermediate 
frame to assure that member proportions are within the scope of the present research experience on 
seismic resistance and that analysis, design procedures, qualities of the materials, and inspection 
procedures are at the highest level of the state of the art. 
9.5 SEISMIC DESIGN CATEGORIES D, E, AND F 
The requirements conform to current practice in the areas of highest seismic hazard. 
9.6 ACCEPTANCE CRITERIA FOR SPECIAL PRECAST STRUCTURAL WALLS 
BASED ON VALIDATION TESTING 
9.6.1. Notation 
Symbols additional to those in Chapter 21 of ACI 318 are defined: 
Ah = area of hysteresis loop 
E1,E2 = peak lateral resistance for positive and negative loading, respectively, for third cycle of loading 
sequence. 
f1 = live load factor defined in 9.6.2.3. 
hw = height of column of test module, in. or mm. 
K, K = initial stiffness for positive and negative loading, respectively, for first cycle 
.1,.2 = drift ratios at peak lateral resistance for positive and negative loading, respectively, for third 
cycle of loading sequence.. 
.1',.2' = drift ratios for zero lateral load for unloading at stiffness K, K from peak positive and 
negative lateral resistance, respectively, for third cycle of loading sequence. (Figure C9.6.2.4) 
. = lateral displacement, in. or mm. See Figures. C9.6.2.2.1, C9.6.2.2.2 and C9.6.2.2.3 
.a = allowable story drift, in. or mm. See Table 9.5.2.8 of SEI/ASCE 7-02 
9.6.2 Definitions 
9.6.2.1 Coupling elements. Coupling elements are connections provided at specific intervals along the 
vertical boundaries of adjacent structural walls. Coupled structural walls are stiffer and stronger than the 
same walls acting independently. For cast-in-place construction effective coupling elements are typically 
coupling beams having small span-to-depth ratios. The inelastic behavior of such beams is normally 
controlled by their shear strength. For precast construction, effective coupling elements can be precast 
beams connected to the adjacent structural walls either by post-tensioning, ductile mechanical devices, or 
grouted-in-place reinforcing bars. The resultant coupled construction can be either emulative of cast-in-
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place construction or non-emulative(jointed). However, for precast construction coupling beams can also 
be omitted and mechanical devices used to connect directly the vertical boundaries of adjacent structural 
walls2,3. 
9.6.2.2 Drift ratio. The definition of the drift ratio, ., is illustrated in Figure C9.6.2.2.1 for a three panel 
wall module. The position of the module at the start of testing, with only its self-weight acting, is 
indicated by broken lines. The module is set on a horizontal foundation support that is centered at A and 
is acted on by a lateral force H applied at the top of the wall. The self-weight of the wall is distributed 
uniformly to the foundation support. However, under lateral loading, that self-weight and any axial 
gravity load acting at the top of the wall cause overturning moments on the wall that are additional to the 
overturning moment Hhw and can affect deformations. The chord AB of the centroidal axis of the wall is 
the vertical reference line for drift measurements. 
For acceptance testing a lateral force H is applied to the wall through the pin at B. Depending on the 
geometric and reinforcement characteristics of the module that force can result in the module taking up 
any one, or a combination, of the deformed shapes indicated by solid lines in Figures C9.6.2.2.1, 
C9.6.2.2.2 and C9.6.2.2.3. 

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Figure C9.6.2.2.2 illustrates several possible components of the displacement . for a wall that is 
effectively solid while Figure C9.6.2.2.3 illustrates two possibly undesirable components of the 
displacement .. Regardless of the mode of deformation of the wall, the lateral force causes the wall at B 
to displace horizontally by an amount .. The drift ratio is the angular rotation of the wall chord with 
respect to the vertical and for the setup shown equals . / hw where hw is the wall height and is equal to the 
distance between the foundation support at A and the load point at B. 
Where prestressing steel is used in wall members, the stress fps in the reinforcement at the nominal and the 
probable lateral resistance shall be calculated in accordance with Sec. 18.7 of ACI 318. 
9.6.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural 
walls based on validation testing. The requirements of Sec. 21.2.1.5 of ACI 318 concerning toughness 
cover both to the energy dissipation of the wall system which, for monolithic construction, is affected 
primarily by local plastic hinging behavior and the toughness of the prototype structure as a whole. The 
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latter is termed global toughness in these provisions and is a condition that does not apply to the walls 
alone. That global toughness requirement can be satisfied only though analysis of the performance of the 
prototype structure as a whole when the walls perform to the criteria specified in these provisions. 
The required gravity load for global toughness evaluations is the value given by these provisions. For 
conformity with Sec. 9.2.1 of ACI 318, UBC 1997, IBC 2003 and NFPA 5000, the required gravity load 
is 1.2D + f1L where the seismic force is additive to gravity forces and 0.9D where the seismic force 
counteracts gravity forces. D is the effect of dead loads, L is the effect of live loads, and f1 is a factor 
equal to 0.5 except for garages, areas occupied as places of public assembly, and all areas where the live 
load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0. 
9.6.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C9.6.2.5 for the third 
loading cycle to the limiting drift ratio required by Sec. 9.6.7.4, 9.6.7.5 or 9.6.7.6, as appropriate. 
For Figure C9.6.2.5, it is assumed that the test module has exhibited different initial stiffnesses, K and 
K, for positive and negative lateral forces and that the peak lateral resistances for the third cycle for the 
positive and negative loading directions, E1 and E2, also differ. The area of the hysteresis loop for the 
third cycle, Ah, is hatched. The circumscribing figure consists of two parallelograms, ABCD and DFGA. 
The slopes of the lines AB and DC are the same as the initial stiffness, K, for positive loading and the 
slopes of the lines DF and GA are the same as the initial stiffness, K', for negative loading. The relative 
energy dissipation ratio concept is similar to the equivalent viscous damping concept used in Sec. 13.9.3 
of the 2000 NEHRP Provisions and Commentary for required tests of seismic isolation systems. 
For a given cycle the relative energy dissipation ratio, , is the area, Ah, inside the lateral force-drift ratio 
loop for the module, divided by the area of the effective circumscribing parallelograms ABCD and 
DFGA. The areas of the parallelograms equal the sum of the absolute values of the lateral force strengths, 
E1 and E2, at the drift ratios .1 and .2 multiplied by the sum of the absolute values for the drift ratios .1' 
and .2'. 
9.6.3 Scope and general requirements. While only ACI Committee 318 can determine the 
requirements necessary for precast walls to meet the provisions of Sec. 21.2.1.5 of ACI 318, Sec. 1.4 of 
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ACI 318 already permits the building official to accept wall systems, other than those explicitly covered 
by Chapter 21 of ACI 318, provided specific tests, load factors, deflection limits, construction procedures 
and other pertinent requirements have been established for acceptance of such systems consistent with the 
intent of the code. The purpose of these provisions is to provide a framework that establishes the specific 
tests, load factors, deflection limits and other pertinent requirements appropriate for acceptance, for 
regions of high seismic risk or for structures assigned to high seismic performance or design categories, 
of precast wall systems, including coupled wall systems, not satisfying all the requirements of Chapter 21 
of ACI 318. For regions of moderate seismic risk or for structures assigned to intermediate seismic 
performance or design categories, less stringent provisions than those specified here are appropriate. 
These provisions assume that the precast wall system to be tested has details differing from those 
prescribed by Sec. 21.7 of ACI 318 for conventional monolithic reinforced concrete construction. Such 
walls may, for example, involve the use of precast elements, precast prestressed elements, post-tensioned 
reinforcement, or combinations of those elements and reinforcement. 
For monolithic reinforced concrete walls a fundamental design requirement of Chapter 21 of ACI 318 is 
that walls with hw/lw exceeding1.0 be proportioned so that their inelastic response is dominated by flexural 
action on a critical section located near the base of the wall. That fundamental requirement is retained in 
these provisions. The reason is that tests on modules, as envisioned in these provisions, cannot be 
extrapolated with confidence to the performance of panelized walls of proportions differing from those 
tested for the development of Chapter 21of ACI 318 if the shear-slip displacement pattern of Figure 
C9.6.2.2.3, or the shear deformation response of Figure C9.6.2.2.2, governs the response developed in the 
test on the module. Two other fundamental requirements of Chapter 21 of ACI 318 are for ties around 
heavily strained boundary element reinforcement and the provision of minimum amounts of uniformly 
distributed horizontal and vertical reinforcement in the web of the wall. Ties around boundary element 
reinforcement to inhibit its buckling in compression are required where the strain in the extreme 
compression fiber is expected to exceed some critical value. Minimum amounts of uniformly distributed 
horizontal and vertical reinforcement over the height and length of the wall are required to restrain the 
opening of inclined cracks and allow the development of the drift ratios specified in Sec. 9.6.7.4, 9.6.7.5 
and 9.6.7.6. Deviations from those tie and distributed reinforcement requirements are possible only if a 
theory is developed that can substantiate reasons for such deviations and that theory is tested as part of 
the validation testing. 
9.6.3.1. These provisions are not intended for use with existing construction or for use with walls that are 
designed to conform to all the requirements of Sec. 21.7 of ACI 318. The criteria of these provisions are 
more stringent than those for walls designed to Sec. 21.7 of ACI 318. Some walls designed to 21.7, and 
having low height to length ratios, may not meet the drift ratio limits of Eq. 9.6.1 because their behavior 
may be governed by shear deformations. The height to length ratio of 0.5 is the least value for which Eq. 
9.6.1 is applicable. 
9.6.3.3. For acceptance, the results of the tests on each module must satisfy the acceptance criteria of 
Sec. 9.6.9. In particular, the relative energy dissipation ratio calculated from the measured results for the 
third cycle between the specified limiting drift ratios must equal or exceed 1/8. For uncoupled walls, 
relative energy dissipation ratios increase as the drift ratio increases. Tests on slender monolithic walls 
have shown relative energy dissipation ratios, derived from rotations at the base of the wall, of about 40- 
45 percent at large drifts. The same result has been reported even where there has been a significant 
opening in the web of the wall on the compression side. For 0.020 drift ratios and walls with height to 
length ratios of 4, relative energy dissipation ratios have been computed as 30, 18, 12, and 6 percent, for 
monolithic reinforced concrete, hybrid reinforced/post-tensioned prestressed concrete with equal flexural 
strengths provided by the prestressed and deformed bar reinforcement, hybrid reinforced/post-tensioned 
prestressed concrete with 25 percent of the flexural strength provided by deformed bar reinforcement and 
75 percent by the prestressed reinforcement, and post-tensioned prestressed concrete special structural 
walls, respectively. Thus, for slender precast uncoupled walls of emulative or non-emulative design it is 
to be anticipated that at least 35 percent of the flexural capacity at the base of the wall needs to be 
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206 
provided by deformed bar reinforcement if the requirement of a relative energy dissipation ratio of 1/8 is 
to be achieved. However, if more than about 40 percent of the flexural capacity at the base of the wall is 
provided by deformed bar reinforcement, then the self-centering capability of the wall following a major 
event is lost and that is one of the prime advantages gained with the use of post-tensioning. For squat 
walls with height to length ratios between 0.35 and 0.69 the relative energy dissipation has been 
reported13 as remaining constant at 23 percent for drifts between that for first diagonal cracking and that 
for a post-peak capacity of 80 percent of the peak capacity. Thus, regardless of whether the behavior of a 
wall is controlled by shear or flexural deformations a minimum relative energy dissipation ratio of 1/8 is a 
realistic requirement. 
For coupled wall systems, theoretical studies and tests have demonstrated that the 1/8 relative energy 
dissipation ratio can be achieved by using central post-tensioning only in the walls and appropriate 
energy dissipating coupling devices connecting adjacent vertical wall boundaries. 
9.6.3.3.4. The SEI/ASCE 7-02 allowable story drift limits are the basis for the drift limits of IBC 2003 
and NFPA 5000. Allowable story drifts, .a, are specified in Table 1617.3 of IBC 2003 and likely values 
are discussed in the Commentary to Sec. 9.6.7.4. The limiting initial drift ratio consistent with .a equals 
.a/fCdhw, where f is the strength reduction factor appropriate to the condition, flexure or shear, that 
controls the design of the test module. For example, for .a/hw equal to 0.015, the required deflection 
amplification factor Cd of 5, and f equal to 0.9, the limiting initial drift ratio, corresponding to B in 
Figure C9.6.9.1, is 0.0033. The use of a f value is necessary because the allowable story drifts of the IBC 
are for the design seismic load effect, E, while the limiting initial drift ratio is at the nominal strength, En , 
which must be greater than E/f. The load-deformation relationship of a wall becomes significantly nonlinear 
before the applied load reaches Ent. While the load at which that non-linearity becomes marked 
depends on the structural characteristics of the wall, the response of most walls remains linear up to about 
75 percent of Ent. 
9.6.3.3.5. The criteria of Sec. 9.6.9 are for the test module. In contrast, the criterion of Sec. 9.6.3.3.5 is 
for the structural system as a whole and can be satisfied only by the philosophy used for the design and 
analysis of the building as a whole. The criterion adopted here is similar to that described in the last 
paragraph of R21.2.1 of ACI 318 and the intent is that test results and analyses demonstrate that the 
structure, after cycling three times through both positive and negative values of the limiting drift ratio 
specified in Sec. 9.6.7.4, 9.6.7.5 or 9.6.7.6, as appropriate, is still capable of supporting the gravity load 
specified as acting on it during the earthquake. 
9.6.4 Design procedure 
9.6.4.1. The test program specified in these provisions is intended to verify an existing design procedure 
for precast structural walls for a specific structure or for prequalifying a generic type of special precast 
wall system for construction in general. The test program is not for the purpose of creating basic 
information on the strength and deformation properties of such systems for design purposes. Thus, the 
test modules should not fail during the validation testing, a result that is the opposite of what is usually 
necessary during testing in the development phase for a new or revised design procedure. For a generic 
precast wall system to be accepted based on these provisions, a rational design procedure is to have been 
developed prior to this validation testing. The design procedure is to be based on a rational consideration 
of material properties and force transfer mechanisms, and its development will usually require 
preliminary and possibly extensive physical testing that is not part of the validation testing. Because 
special wall systems are likely to respond inelastically during design-level ground shaking, the design 
procedure must consider wall configuration, equilibrium of forces, compatibility of deformations, the 
magnitudes of the lateral drifts, reversed cyclic displacements, the relative values of each limiting 
engineering design criteria (shear, flexure and axial load) and use appropriate constitutive laws for 
materials that include considerations of effects of cracking, loading reversals and inelasticity. 
The effective initial stiffness of the structural walls is important for calculating the fundamental period of 
the prototype structure. The procedure used to determine the effective initial stiffness of the walls is to be 
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verified from the validation test results as described in Sec. 9.6.7.11. 
Provisions Sec. 9.6.4.1.1 through 9.6.4.1.3 state the minimum procedures to be specified in the design 
procedure prior to the start of testing. The Authority Having Jurisdiction may require that more details be 
provided in the design procedure than those of Sec. 9.6.4.1.1 through 9.6.4.1.3 prior to the start of testing. 
9.6.4.2. The justification for the small number of test modules, specified in Sec. 9.6.5.1 is that a previously 
developed rational design procedure is being validated by the test results. Thus, the test modules for the 
experimental program must be designed using the procedure intended for the prototype wall system and strengths 
must be predicted for the test modules before the validation testing is started. 
9.6.5 Test modules. 
9.6.5.1. One module must be tested for each limiting engineering design criterion, such as shear, or axial 
load and flexure, for each characteristic configuration of walls. Thus, in accordance with 9.6.4.3 if the 
test on the module results in a maximum shear stress of 3 c f ' then the maximum shear stress that can 
be used in the prototype is that same value. Each characteristic in-plane configuration of walls, or coupled 
walls, in the prototype structure must also be tested. Thus, as a minimum for one-way structural walls, 
two modules with the configuration shown in Figure C9.6.2.2.1, and, for one way coupled walls, two 
modules with the configuration shown in either Figure C9.6.5.1(a) or in Figure C9.6.5.1(b), must be 
tested. In addition, if intersecting wall systems are to be used then the response of the wall systems for 
the two orthogonal directions needs to be tested. For two-way wall systems and coupled wall-frame 
systems, testing of configurations other than those shown in Figures C9.6.2.2.1 and C9.6.5.1 may be 
appropriate when it is difficult to realistically model the likely dominant earthquake deformations using 
orthogonal direction testing only. 
This provision should not be interpreted as implying that only two tests will need to be made to qualify a 
generic system. During the development of that system it is likely that several more tests will have been 
made, resulting in progressive refinements of the mathematical model used to describe the likely 
performance of the generic structural wall system and its construction details. Consequently, only one 
test of each module type for each limiting engineering design condition, at a specified minimum scale and 
subjected to specific loading actions, may be required to validate the system. Further, as stated in Sec. 
9.6.9.1, if any one of those modules for the generic wall system fails to pass the validation testing 
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required by these provisions, then the generic wall system has failed the validation testing 
In most prototype structures, a slab is usually attached to the wall and, as demonstrated by the results of 
the PRESSS building test, the manner in which the slab is connected to the wall needs to be carefully 
considered. The connection needs to be adequate to allow the development of story drifts equal to those 
anticipated in these provisions. However, in conformity with common practice for the sub-assemblage 
tests used to develop the provisions of Chapter 21of ACI 318, there is no requirement for a slab to be 
attached to the wall of the test module. The effect of the presence of the slab should be examined in the 
development program that precedes the validation testing. 
9.6.5.3. Test modules need not be as large as the corresponding walls in the prototype structure. The 
scale of the test modules, however, must be large enough to capture all the complexities associated with 
the materials of the prototype wall, its geometry and reinforcing details, load transfer mechanisms, and 
joint locations. For modules involving the use of precast elements, for example, scale effects for load 
transfer through mechanical connections should be of particular concern. The issue of the scale 
necessary to capture fully the effects of details on the behavior of the prototype should be examined in the 
development program that precedes the validation testing. 
9.6.5.4. It is to be expected that for a given generic precast wall structure, such as an unbonded centrally 
post-tensioned wall constructed using multiple precast or precast pretensioned concrete wall panels, 
validation testing programs will initially use specific values for the specified strength of the concrete and 
reinforcement in the walls, the layout of the connections between panels, the location of the posttensioning, 
the location of the panel joints, and the design stresses in the wall. Pending the development 
of an industry standard for the design of such walls, similar to the standard for special hybrid moment 
frames, specified concrete strengths, connection layouts, post-tensioning amounts and locations, etc., used 
for such walls will need to be limited to the values and layouts used in the validation testing programs. 
9.6.5.5. For walls constructed using precast or precast/prestressed panels and designed using nonemulative 
methods, the response under lateral load can change significantly with joint opening (Figure 
C9.6.2.2.2(d) and Figure C9.6.2.2.3(a)). The number of panels used to construct a wall depends on wall 
height and design philosophy. If, in the prototype structure, there is a possibility of horizontal joint 
opening under lateral loading at a location other than the base of the wall, then the consequences of that 
possibility need to be considered in the development and validation test programs. Joint opening at 
locations other than the base can be prevented through the use of capacity design procedures. 
9.6.5.6. The significance of the magnitude of the gravity load that acts simultaneously with the lateral 
load needs to be addressed during the validation testing if the development program suggests that effect is 
significant. 
9.6.5.7. Details of the connection of walls to the foundation are critical, particularly for non-emulative 
wall designs. The deformations that occur at the base of the wall due to plastic hinging or extension of 
the reinforcing bars or post-tensioning steel crossing the wall to foundation interface, (Figure 
C9.6.2.2.2(d)), are in part determined by details of the anchorage and the bonding of those reinforcements 
on either side of the interface. Grout will be normally used to bed panels on the foundation and the 
characteristics of that grout in terms of materials, strength and thickness, can have a large effect on wall 
performance. The typical grout pad with a thickness of 1 inch (25 mm) or less can be expected to provide 
a coefficient of friction of about 0.6 under reversed loadings. Pads with greater thickness and without 
fiber reinforcement exhibit lesser coefficients of friction. Adequate frictional resistance is essential to 
preventing undesirable shear-slip deformations of the type shown in Figure C9.6.2.2.3(b). 
9.6.5.8. The geometry of the foundations need not duplicate that used in the prototype structure. 
However, the geometric characteristics of the foundations (width, depth and length) need to be large 
enough that they do not influence the behavior of the test module. 
9.6.6 Testing agency. In accordance with the spirit of the requirements of Sec.1.3.5 and 1.4 of ACI 318, 
it is important that testing be carried out by a recognized independent testing agency, approved by the 
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agency having jurisdiction and that the testing and reporting be supervised by a registered design 
professional familiar with the proposed design procedure and experienced in testing and seismic 
structural design. 
9.6.7 Test method. The test sequence is expressed in terms of drift ratio, and the initial ratio is related to 
the likely range of linear elastic response for the module. That approach, rather than testing at specific 
drift ratios of 0.005, 0.010, etc., is specified because, for modules involving prestressed concrete, the 
likely range of elastic behavior varies with the prestress level. 
An example of the test sequence specified in Sec. 9.6.7.2 through 9.6.7.6 is illustrated in Figure C9.6.7. 
The sequence is intended to ensure that displacements are increased gradually in steps that are neither too 
large nor too small. If steps are too large, the drift capacity of the system may not be determined with 
sufficient accuracy. If the steps are too small, the system may be unrealistically softened by loading 
repetitions, resulting in artificially low maximum lateral resistances and artificially high maximum drifts. 
Also, when steps are too small, the rate of change of energy stored in the system may be too small 
compared with the change occurring during a major event. Results, using such small steps, can mask 
undesirable brittle failure modes that might occur in the inelastic response range during a major event. 
Because significant diagonal cracking is to be expected in the inelastic range in the web of walls, and in 
particular in squat walls, the pattern of increasing drifts used in the test sequence can markedly affect 
diagonal crack response in the post-peak range of behavior. 
The drift capacity of a building in a major event is not a single quantity, but depends on how that event 
shakes the structure. In the forward near field, a single pulse may determine the maximum drift demand, 
in which case a single large drift demand cycle for the test module would give the best estimation of the 
drift capacity. More often, however, many small cycles precede the main shock and that is the scenario 
represented by the specified loading. 
There is no requirement for an axial load to be applied to the wall simultaneously with the application of 
the lateral displacements. In many cases it will be conservative not to apply axial load because, in 
general, the shear capacity of the wall and the resistance to slip at the base of the wall increase as the axial 
load on the wall increases. However, as the height of the wall increases and the limiting drift utilized in 
the design of the wall increases, the likelihood of extreme fiber crushing in compression at maximum 
drift increases, and the importance of the level of axial load increases. The significance of the level of 
2003 Commentary, Chapter 9 
210 
axial loading should be examined during the development phase. 
9.6.7.4. For the response of a structure to the design seismic shear force, current building codes such as 
UBC-97, IBC 2003 or NFPA 5000 , or recommended provisions such as 2000 Provisions, SEI/ASCE 7- 
02 and FEMA 273 specify a maximum allowable drift. However, structures designed to meet that drift 
limit may experience greater drifts during an earthquake equal to the design basis earthquake and are 
likely to experience greater drifts during an earthquake equal to the maximum credible earthquake. In 
addition to the characteristics of the ground motion, actual drifts will depend on the strength of the 
structure, its initial elastic stiffness, and the ductility expected for the given lateral load resisting system. 
Specification of suitable limiting drifts for the test modules requires interpretation and allowance for 
uncertainties in the assumed ground motions and structural properties. 
In IBC 2003, the design seismic shear force applied at the base of a building is related directly to its 
weight and the design elastic response acceleration, and inversely to a response modification factor, R. 
That R factor increases with the expected ductility of the lateral force resisting system of the building. 
Special structural walls satisfying the requirements of Sec. 21.2 and 21.7 are assigned an R value of 6 
when used in a building frame system and a value of 5 when used in a bearing wall system. They are also 
assigned allowable story drift ratios that are dependent on the hazard to which the building is exposed. 
When the design seismic shear force is applied to a building, the building responds inelastically and the 
resultant computed drifts, (the design story drifts), must be less than a specified allowable drift. 
Additional guidance is given in FEMA 356 where the deformations for rectangular walls with height to 
length ratios greater than 2.5, and flanged wall sections with height to length ratios greater than 3.5, are to 
be assumed to be controlled by flexural actions. When structural walls are part of a building representing 
a substantial hazard to human life in the event of a failure, the allowable story drift ratio for shear 
controlled walls is 0.0075 and for flexure controlled walls is a function of the plastic hinge rotation at the 
base of the wall. For flexure controlled walls values range up to a maximum of about 0.02 for walls with 
confined boundary elements with low reinforcement ratios and shear stress less than 3 c f ' 
To compensate for the use of the R value, IBC Sec. 1617.4.6 requires that the drift determined by an 
elastic analysis for the code-prescribed seismic forces be multiplied by a deflection amplification factor, 
Cd ,to determine the design story drift and that the design story drift must be less than the allowable story 
drift. In building frame systems, structural walls satisfying the requirements of Sec. 21.7of ACI 318 are 
assigned a Cd value of 5. However, research8 has found that design story drift ratios determined in the 
foregoing manner may be too low. Drift ratios of 6 times IBC-calculated values, (rather than 5), are more 
representative of the upper bounds to expected drift ratios. The value of 6 is also in agreement with the 
finding that the drift ratio of an inelastic structure is approximately the same as that of an elastic structure 
with the same initial period. For flexure controlled walls the value of 6/5 times the present IBC limits on 
calculated drift ratio, would lead to a limit on real drift ratios of up to 0.024. 
Duffy et al. reviewed experimental data for shear walls to define post-peak behavior and limiting drift 
ratios for walls with height to length ratios between 0.25 and 3.5. Seo et al. re-analyzed the data of Duffy 
et al. together with data from tests conducted subsequent to the analysis of Duffy et al. Duffy et al. 
established that for squat walls with web reinforcement satisfying ACI 318-02 requirements and height to 
length ratios between 0.25 and 1.1, there was a significant range of behavior for which drifts were still 
reliable in the post-peak response region. Typically the post-peak drift increased by 0.005 for a 20 
percent degradation in capacity under cyclic loading. For greater values of degradation, drifts were less 
reliable. That finding has also been confirmed through tests conducted by Hidalgo et al.13 on squat walls 
with effective height to length ratios ranging between 0.35 and 1.0. Values of the drift ratio of the walls 
at inclined cracking and at peak capacity varied little with web reinforcement. By contrast, drifts in the 
post-peak range were reliable to a capacity equal to 80 percent of the peak capacity and were 0.005 
greater than the drifts at peak capacity provided the walls contained horizontal and vertical web 
reinforcement equal to 0.25 percent. 
From an analysis of the available test data, and from theoretical considerations for a wall rotating 
Concrete Structure Design Requirements 
211 
flexurally about a plastic hinge at its base, Seo et al concluded that the limiting drift at peak capacity 
increased almost linearly with the height to length ratio of the wall. When the additional post peak drift 
capacity for walls with adequate web reinforcement was added to the drift at peak capacity, then the total 
available drift capacity in percent was given by the following equation: 
1.0=0.67[hw lw]+0.5=3.0 
where hw is the height of the wall, and lw is the length of the wall. The data from the tests of Hidalgo et 
al. suggest that while that formula is correct for squat walls the lower limit on drift can be decreased to 
0.8 as specified in these provisions and that the use of that formula should be limited to walls with height 
to length ratios equal to or greater than 0.5. For wall height to length ratios less than 0.5 the behavior is 
controlled principally by shear deformations, (Figure C9.6.2.2.2(c)), and Eq. 9.6.1 should not be used. 
The upper value of 0.030 for the drift ratio was somewhat optimistic because the data were for walls with 
height to length ratios equal to or less than 3.5 and subsequent tests have shown that the upper limit of 
2.5, as specified in Eq. 9.6.1, is a more realistic limit. 
9.6.7.5. The design capacity for coupled wall systems must be developed by the drift ratio corresponding 
to that for the wall with the least hw/lw value. However, it is desirable that testing be continued to the drift 
given by Eq. 9.6.1 for the wall with the greatest hw/lw in order to assess the reserve capacity of the coupled 
wall system. 
9.6.7.6. The drift limits of Eq. 9.6.1 are representative of the maximum that can be achieved by walls 
designed to ACI 318. The use of smaller drift limits is appropriate if the designer wishes to use 
performance measures less than the maximum permitted by ACI 318. Examples are the use of reduced 
shear stresses so that the likelihood of diagonal cracking of the wall is minimized or reduced compressive 
stresses in the boundary elements of the wall so that the risk of crushing is reduced. Non-linear time 
history analyses for the response to a suite of maximum considered earthquake (MCE) ground motions, 
rather than 1.5 times a suite of the corresponding design basis earthquake (DBE) ground motions, is 
required because the drifts for the response to the MCE motion can be significantly larger than 1.5 times 
the drifts for the response to the DBE motions. 
9.6.7.10. In many cases, data additional to the minimum specified in Sec. 9.6.7.7 may be useful to 
confirm both design assumptions and satisfactory response. Such data include relative displacements, 
rotations, curvatures, and strains. 
9.6.8 Test report. 
The test report must be sufficiently complete and self-contained for a qualified expert to be satisfied that 
the tests have been designed and carried out in accordance with these criteria, and that the results satisfy 
the intent of these provisions. Sec.9.6.8.1.1 through 9.6.8.1.11 state the minimum evidence to be 
contained within the test report. The Authority having Jurisdiction or the registered design professional 
supervising the testing may require that additional test information be reported. 
9.6.9 Test module acceptance criteria. 
The requirements of this clause apply to each module of the test program and not to an average of the 
results of the program. Figure C9.6.9.1 illustrates the intent of this clause. 
2003 Commentary, Chapter 9 
212 
9.6.9.1.1. Where nominal strengths for opposite loading directions differ, as is likely for C-, L- or Tshaped 
walls, the criterion of Sec. 9.6.9.1.1 applies separately to each direction. 
9.6.9.1.2. At high cyclic-drift ratios, strength degradation is inevitable. To limit the level of degradation 
so that drift ratio demands do not exceed anticipated levels, a maximum strength degradation of 0.20Emax 
is specified. Where strengths differ for opposite loading directions, this requirement applies 
independently to each direction. 
9.6.9.1.3. If the relative energy dissipation ratio is less than 1/8, there may be inadequate damping for the 
building as a whole. Oscillations may continue for some time after an earthquake, producing low-cycle 
fatigue effects, and displacements may become excessive. 
If the stiffness becomes too small around zero drift ratio, the structure will be prone to large 
displacements for small lateral force changes following a major earthquake. A hysteresis loop for the 
third cycle between peak drift ratios of 1/10 times the limiting drift ratio given by Eq. 9.6.1, that has the 
form shown in Figure C9.6.9.1, is acceptable. At zero drift ratio, the stiffnesses for positive and negative 
loading are about 11 percent of the initial stiffnesses. Those values satisfy Sec. 9.6.9.1.3. An 
unacceptable hysteresis loop form would be that shown in Figure C9.6.9.1.3 where the stiffness around 
zero drift ratio is unacceptably small for both positive and negative loading. 

Concrete Structure Design Requirements 
213 
Figure C 9.6.9.1.3 Unacceptable hysteretic behavior 
REFERENCES 
ACI Innovation Task Group I and Collaborators. 2001. Acceptance Criteria for Moment Frames Based on 
Structural Testing (T1.1-01) and Commentary(T1.1R-01). American Concrete Institute, Farmington Hills, 
MI, 2001, 10p. 
ACI Innovation Task Group I and Collaborators. 2001  Special Hybrid Moment Frames Composed of 
Discretely Jointed Precast and Post-Tensioned Concrete Members (ACI T1.2-XX) and Commentary (ACI 
T1.2R-XX), ACI Structural Journal,.98 (5):771-784. 
Ali, A. and J.K. Wight. 1990. Reinforced Concrete Structural Walls with Staggered Opening 
Configurations Under Reversed Cyclic Loading, Report UMCE 90-05, Department of Civil Engineering, 
University of Michigan, Ann Arbor, MI. 
Building Seismic Safety Council. 1987. Guide to Use of NEHRP Recommended Provisions in 
Earthquake Resistant Design of Buildings, 1985 Edition. Washington, D.C.: FEMA. 
Building Seismic Safety Council, NEHRP Recommended Provisions for Seismic Regulations for New 
Buildings and Other Structures, 2000. FEMA 368 and 369. Washington, D.C. 
Cheok, G. S., and H. S. Lew. 1991. Performance of Precast Concrete Beam-to-Column Connections 
Subject to Cyclic Loading, PCI Journal, 36, (3):56-67. 
Duffy, T.A., A.Goldman, and C.R. Farrar. 1993. Shear Wall Ultimate Drift Limits, Report 
NUREG/CR-6104, LA-12649-MS. U.S. Nuclear Regularity Commission. 
Elliott, K. S., G. Davies, and W. Omar. 1992. Experimental and Theoretical Investigation of Precast 
Concrete Hollow-Cored Slabs Used as Horizontal Floor Diaphragms, The Structural Engineer, 70, 
(10):175-187. 
Englekirk, R. E. 1987. Concepts for the Development of Earthquake Resistant Ductile Frames of 
Precast Concrete, PCI Journal, 32, (1). 
Federal Emergency Management Agency. 2000. NEHRP Guidelines for the Seismic Rehabilitation of 
2 003 Commentary, Chapter 9 
214 
Buildings, Chapter 6, Concrete, FEMA 356 and 357. 
French, C. W., M. Hafner, and V. Jayashanker. 1989. Connections Between Precast Elements - Failure 
Within Connection Region, ASCE Journal of Structural Engineering, 115, (12):3171-3192. 
Hawkins, N. M., and R. E. Englekirk. 1987. U.S.-Japan Seminar on P/C Concrete Construction in 
Seismic Zones, PCI Journal, 32,(2). 
Hidalgo, P.A.,C.A. Ledezma, and R.A. Jordan. 2002. Seismic Behavior of Squat Reinforced Concrete 
Shear Walls, Earthquake Spectra, 18 (2): 287-308. 
Hutchinson, R.L., S.H. Rizkalla, M. Lau, and M. Heuvel, 1991. Horizontal Post-Tensioned Connections 
for Precast Concrete Bearing Shear Walls, PCI Journal, 36,(3):.64-76. 
International Code Council. 2003. International Building Code, Falls Church, VA. 
International Conference of Building Officials, Uniform Building Code. 1997, Vol. 2, Structural 
Engineering Design Provisions. Whittier, CA. 
Ishizuka, T., and N. M. Hawkins. 1987. Effect of Bond Deterioration on the Seismic Response of 
Reinforced and Partially Prestressed Concrete and Ductile Moment Resistant Frames, SM87-2, 
University of Washington, Department of Civil Engineering. 
Jayashanker, V., and C. E. French. 1988. An Interior Moment Resistant Connection Between Precast 
Elements Subjected to Cyclic Lateral Loads, Structural Engineering Report No. 87-10, Minneapolis: 
University of Minnesota. 
Kurama, Y.C. 2002. Hybrid Post-Tensioned Precast Concrete Walls for Use in Seismic Regions. PCI 
Journal, 47, (5): 36-59. 
Mast, R. F. 1992. A Precast Concrete Frame System for Seismic Zone Four, PCI Journal 37(1):50-64. 
Nakaki, S. D., and R. E. Englekirk. 1991. PRESSS Industry Seismic Workshops: Concept 
Development, PCI Journal 36(5):54-61. 
National Fire Protection Association . 2003. Building Construction and Safety Code, NFPA 5000, 
Quincy, MA. 
Neille, D. S. 1977. Behavior of Headed Stud Connections for Precast Concrete Connections for Precast 
Concrete Panels Under Monotonic and Cycled Shear Loading, thesis submitted in partial fulfillment of 
the requirements of Doctor of Philosophy, University of British Columbia. 
New Zealand Society for Earthquake Engineering. 1991. Guidelines for the Use of Structural Precast 
Concrete in Buildings. 
Park, R., and K. J. Thompson. 1977. Cyclic Load Tests on Prestressed and Partially Prestressed 
Beam-Column Joints, PCI Journal 22(5):84-110. 
PCI Ad Hoc Committee on Precast Walls. 1997. Design for Lateral Force Resistance with Precast 
Concrete Shear Walls, PCI Journal, 42, (2): 44-65. 
Pekau, O. A., and D. Hum. 1991. Seismic Response of Friction-Jointed Precast Panel Shear Walls, 
PCI Journal 36(2):56-71. 
Powell, G., F. Filippou, V. Prakash, and S. Campbell. 1993. Analytical Platform for Precast Structural 
Systems, In Proceedings, ASCE Structures Congress '93. New York: ASCE. 
Priestley, M. J. N. 1991. Overview of PRESSS Research Program, PCI Journal 36(4):50-57. 
Priestley, M. J. N., and J. T. Tao. 1993. Seismic Response of Precast Prestressed Concrete Frames with 
Partially Debonded Tendons, PCI Journal 38(1):58-69. 
Priestley, M.J.N., S. Sritharan, J. Conley, and S. Pampanin. 1999. Preliminary Results and Conclusions 
 Concrete Structure Design Requirements 
215 
from the PRESSS Five-Story Precast Concrete Test Building, PCI Journal, 44,(6): 42-67. 
Schultz, A.E. and R.A. Magana. 1996. Seismic Behavior of Connections in Precast Concrete Walls, 
Paper SP 162-12. Mete A. Sozen Symposium, pp.273-311, ACI SP 162. American Concrete Institute, 
Farmington Hills, MI. 
Seo, S-Y., L-H Lee, and N.M. Hawkins. 1998.  The Limiting Drift and Energy Dissipation Ratio for 
Shear Walls Based on Structural Testing, Journal of the Korean Concrete Institute,10, (6): 335-343. 
Stanton, J. F., T. R. Hicks, and N. M. Hawkins. 1991. PRESSS Project 1.3: Connection Classification 
and Evaluation, PCI Journal 36(5):62-71. 
Stanton, J.F. and S.D. Nakaki, 2002. Design Guidelines for Precast Concrete Seismic Structural 
Systems-Unbonded Post-Tensioned Split Walls PRESSS Report 01/03-09, UW Report SM 02-02, 
Department of Civil Engineering, University of Washington, Seattle, WA.. 
Taylor, C.P., P.E. Cote, and J.W. Wallace.  Design of Slender Reinforced Concrete Walls with 
Openings, ACI Structural Journal, 95, (4): 420-433. 
Thompson, K. J., and R. Park. 1980. Seismic Response of Partially Prestressed Concrete, ASCE 
Journal of the Structural Division, 106(ST8):1755-1775. 
Uang, C-M., and A. Maarouf.1993. Seismic Displacement Amplification Factor in Uniform Building 
Code, SEAONC Research Bulletin Board, BB93-3, pp. B1-B2, and Displacement Amplification 
Factor for Seismic Design Provisions, Proc. Structures Congress, ASCE, Vol.1, pp. 211-216. Irvine, CA 
2 003 Commentary, Chapter 9 
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217 
Appendix to Chapter 9 
UNTOPPED PRECAST DIAPHRAGMS 
Although not directly addressed in the code, untopped precast components have been used as 
diaphragms in high seismic regions. Untopped hollow-core planks with grouted joints and end chords 
have performed successfully both in earthquakes and in laboratory tests, (Elliot et al., 1992; 
Menegotto, 1994; Priestley et al., 1999). Experience has also demonstrated the unsuccessful use of 
cast-in-place concrete topping as diaphragms (Iverson and Hawkins, 1994). Where problems have 
occurred, they have not been inherently with the precast construction, but the result of a failure to 
address fundamental requirements of structural mechanics. 
This section provides conditions that are intended to ensure that diaphragms composed of precast 
components are designed with attention to the principles required for satisfactory behavior. Each 
condition addresses requirements that should be considered for all diaphragms, but which are 
particularly important in jointed construction. Specific attention should be paid to providing a 
complete load path that considers force transfer across all joints and connections. 
A9.2 DESIGN REQUIREMENTS 
A9.2.1 Configuration. Out-of-plane offsets in the vertical elements of the seismic-force-resisting 
system place particularly high demands on the diaphragm in providing a continuous load path. 
Untopped precast diaphragms are not suitable for this condition. It must be recognized that the 
demand on diaphragms in buildings with these plan irregularities requires special attention. In 
accordance with Sec. 4.6.3.2 the design force for the diaphragm should be increased by at least 25 
percent when such irregularities are present in structures assigned to Seismic Design Category D, E, or 
F. 
A9.2.2 Diaphragm demand. Following the principle that the diaphragm is not generally an 
appropriate location for inelastic behavior and, in particular, for untopped precast diaphragms, specific 
direction is provided that elastic models should be used for diaphragm analysis. Connections are 
subject to a combination of load effects (Fleischman et al., 1998). The distribution of loads may 
change after yielding, and therefore the design of the diaphragm should avoid yielding. 
Since the diaphragm is not generally an appropriate location for inelastic behavior, it should be 
designed to a level of strength that is intended to ensure that the ductility and yield strength of the 
seismic-force-resisting system can be mobilized before the diaphragm yields. While research 
(Fleischman et al., 1998) suggests that the diaphragm demand will not exceed twice the equivalent 
lateral forces used for the vertical system design, Table 4.3-1 prescribes an overstrength factor, O0
, 
and Sec. 4.3.3 prescribes a redundancy factor, ., for the systems that should be used. If an analysis of 
the probable strength of the seismic-force-resisting system is made to determine a lower demand on 
the diaphragm, the design force used should still be sufficient to attempt to ensure that the diaphragm 
remains elastic. For that reason a 1.25 factor is specified. 
A9.2.3 Mechanical connections. Although the design procedures prescribed in these sections are 
intended to ensure elastic behavior at the level of the code design forces, it is recognized that 
catastrophic events may exceed code requirements. Under such circumstances, it is important that the 
connections possess ductility under reversed cyclic loading. The intent, in these sections, is for the 
connection capacity to be limited by steel yielding of the connector and not by brittle concrete failure 
or weld fracture. 
Substantiating experimental evidence to demonstrate through testing and evaluation that mechanical 
connections satisfy the principles specified in ACI T1.1-01 and ATC-24, and can develop the required 
capacity and ductility, should meet the following criteria: 
2 003 Commentary, Appendix to Chapter 9 
218 
Test Procedures: 
1. Prior to testing, a design procedure should have been developed for prototype connections having 
the generic form that is to be tested for acceptance. 
2. That design procedure should be used to proportion the test specimens. 
3. Specimens should not be less than two-thirds scale. 
4. Test specimens should be subject to a sequence of reversing cycles having increasing limiting 
displacements. 
5. Three fully reversed cycles should be applied at each limiting displacement. 
6. The maximum load for the first sequence of three cycles should be 75 percent of the calculated 
nominal strength of the connection, En. 
7. The stiffness of the connection should be defined as 75 percent of the calculated nominal strength 
of the connection divided by the corresponding measured displacement, dm. 
8. Subsequent to the first sequence of three cycles, limiting displacements should be incremented by 
values not less than 1.0, and not more than 1.25 times dm. 
Acceptance Criteria: 
1. The connection should develop a strength, Emax, greater than its calculated nominal strength, En. 
2. The strength, Emax, should be developed at a displacement not greater than 3dm. 
3. For cycling between limiting displacements not less than 3dm, the peak force for the third loading 
cycle for a given loading direction should not be less than 0.8 Emax for the same loading direction. 
Results of reversed cyclic loading tests on typical connections are reported in Spencer (1986) and 
Pincheira et al. (1998). 
A9.2.4 Cast-in-place strips. Successful designs may include a combination of untopped precast 
components with areas of concrete topping in locations of high force demand or concentration. Such 
topping can allow for continuity of reinforcement across joints. For such designs, the requirements for 
topping slab diaphragms apply to the topped portions. 
A9.2.5 Deformation compatibility. An important element in the Provisions is attention to 
deformation compatibility requirements. Reduction in effective shear and flexural stiffness for the 
diaphragm is appropriate in evaluating the overall effects of drift on elements that are not part of the 
seismic-force-resisting system. This approach should encourage the use of more vertical elements to 
achieve shorter spans in the diaphragm and result in improved system redundancy and diaphragm 
continuity. Redundancy will also improve the overall behavior should any part of the diaphragm yield 
in a catastrophic event. 
REFERENCES 
Elliot, K. S., G. Davies, and W. Omar. 1992. Experimental and Theoretical Investigation of Precast 
Concrete Hollow-cored Slabs Used as Horizontal Floor Diaphragms. The Structural Engineer, 70 
(10) 175-187. 
Fleischman, R. B., R. Sause, S. Pessiki, and A. B. Rhodes. 1998. Seismic Behavior of Precast 
Parking Structure Diaphragms. PCI Journal, 43 (1): 38-53. 
Iverson, J. K., and N. M. Hawkins. 1994. Performance of Precast/Prestressed Concrete Building 
Structures During the Northridge Earthquake, PCI Journal 39(2): 38-55. 
Menegotto, M. 1994. Seismic Diaphragm Behavior of Untopped Hollow-Core Floors, 
Proceedings, F.I.P. Congress, Washington, D.C. 
Untopped Precast Diaphragms 
219 
Pincheira, J. A., M. G. Oliva, and F. I. Kusumo-Rahardjo. 1998. Tests on Double Tee Flange 
Connectors Subject to Monotonic and Cyclic Loading, PCI Journal, 43 (3): 82-96. 
Priestley, M. J. N., S. Sritharan, J. R. Conley, and S. Pampanin. 1999. Preliminary Results and 
Conclusions from the PRESSS Five-Story Precast Concrete Test Building. PCI Journal, 44 (6). 
Spencer, R. 1986. Earthquake Resistant Connections for Low Rise Precast Concrete Buildings, 
Proceedings, U.S.-Japan Seminar on Precast Concrete Construction in Seismic Zones, Vol. 1, Japan 
Concrete Institute. 
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 223 
Chapter 11 Commentary 
MASONRY STRUCTURE DESIGN REQUIREMENTS 
11.1.2 References. The main concern of Chapter 11 is the proper detailing of masonry construction for 
earthquake resistance. The bulk of the detailing requirements in this chapter are contained in ACI 
530/ASCE 5/TMS 402. The commentary for ACI 530/ASCE 5/TMS 402 contains a valuable discussion 
of the rationale behind detailing requirements that is not repeated here. 
11.2.1.5.1 Shear keys. Shear keys provide resistance to the movement of shear walls when yielding of 
the reinforcing steel occurs. This phenomenon was observed in tests by Klingner. (Leiva and Klingner 
1991). There has been no field verification of shear wall movement under seismic events. The shear 
key requirements are based on judgment and sizes are based on current construction procedures. 
11.2.2.3 Article 1.3 permits the use of structural clay wall-tile meeting the requirements of ASTM C 34. 
At the time of publication, it was felt that the existing detailing requirements for masonry elements did 
not adequately address the brittle nature of clay wall-tile units. 
11.2.2.11 The nominal shear strength of coupling beams must be equal to the shear caused by 
development of a full yield hinge at each end of the coupling beams. This nominal shear strength is 
estimated by dividing the sum of the calculated yield moment capacity of each end of the coupling 
beams, M1 and M2, by the clear span length, L. 
A coupling beam may consist of a masonry beam and a part of the reinforced concrete floor system. 
Reinforcement in the floor system parallel to the coupling beam should be considered as a part of the 
coupling beam reinforcement. The limit of the minimum width of floor that should be used is six times 
the floor slab thickness. This quantity of reinforcement may exceed the limits of Sec. 3.2.3.5 but should 
be used for the computation of the normal shear strength. 
11.2.2.12 The theory used for design of beams has a limited applicability to deep beams. Shear warping 
of the cross section and a combination of diagonal tension stress and flexural tension stress in the body 
of the deep beam requires that deep beam theory be used for design of members that exceed the 
specified limits of span to depth ratio. Analysis of wall sections that are used as beams generally will 
result in a distribution of tensile stress that requires the lower one-half of the beam section to have 
uniformly distributed reinforcement. The uniform distribution of reinforcement resists tensile stress 
caused by shear as well as flexural moment. 
The flexural reinforcement for deep beams must meet or exceed the minimum flexural reinforcement 
ratio of Sec. 3.2.4.3.2. Additionally, horizontal and vertical reinforcement must be distributed 
throughout the length and depth of deep beams and must provide reinforcement ratios of at least 
0.0007bd. Distributed flexural reinforcement may be included in the calculations of the minimum 
distributed reinforcement ratios. 
11.2.2.13 Corrugated sheet metal ties are prohibited from use in Seismic Design Categories E and F due 
to their decreased capacity in transferring loads. 
11.2.2.14 Masonry pryout refers to a failure mode of a shear anchor in which the embedded end of the 
anchor moves opposite to the direction of applied shear, prying out a roughly semi-conical body of 
masonry (concrete, as applicable) behind the anchor. It is not the same as a breakout, which refers to 
a failure mode of a shear anchor in which a body of masonry (or concrete, as applicable) is broken off 
between the anchor and a free edge, in the direction of applied shear. 

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224 
11.4 GLASS-UNIT MASONRY AND MASONRY VENEER 
Chapters 11 and 12 of ACI 530-95/ASCE 5-95/TMS 402-95 were introduced into the 1997 Provisions 
to address design of glass-unit masonry and masonry veneer. Direct reference is made to these chapters 
for design requirements. Investigations of seismic performance have shown that architectural 
components meeting these requirements perform well (Jalil, Kelm, and Klingner, 1992; and Klingner, 
1994). 
11.5 PRESTRESSED MASONRY 
Allowable stress provisions are set forth in MSJC Chapter 4. There are no strength design provisions for 
prestressed masonry. There is a paucity of data on the cyclic testing of prestressed shear walls. There is 
only one published report of cyclic testing of prestressed shear walls in-plane using a testing protocol 
similar to the sequential phased displacement method used in the TCCMaR program. This report 
considers specimens both partially and fully grouted using only prestressed bar reinforcing. There is no 
published in-plane cyclic test data using prestressed strand, nor any published data using prestressed 
reinforcing in combination with mild steel reinforcing. There is some additional unpublished data on inplane 
testing of prestressed masonry shear walls using prestressed bars only. 
The data shows that solid grouted prestressed masonry shear walls subjected to in-plane cyclic 
displacements perform as an essentially elastic system with stiffness degradation in each cycle. Little 
energy is dissipated in the hysterisis loops. Although reasonably large displacements can be reached, 
there is essentially no ductile behavior. The data on partially grouted walls is sparse and shows inability 
to reach large displacement before failure. The data shows that MSJC Eq. 3-21 provides a reasonable 
estimate for the shear capacity for solid grouted walls. 
The TCCMaR research showed that the ductility of a masonry wall loaded in-plane was highly 
dependent on the level of axial load and the amount of reinforcing. Ductile behavior declines 
significantly at axial loads in excess of 100 psi; ductile behavior also declines significantly when the 
reinforcement ratio is high. The addition of prestressing to a wall with mild steel reinforcing will 
decrease the ductility. 
Because of the limited data and the potential for non-ductile, prestressed masonry shear walls are 
restricted to Seismic Design Categories A and B and the R factor is set at 1. As more research 
becomes available, these restrictions could be eased. 
11.6 ANCHORING TO MASONRY 
This section covers cast-in-place headed anchor bolts and bent-bar anchors (J- or L-bolts) in grout. 
General background information on this topic is given in CEB, 1995. 
The tensile capacity of a headed anchor bolt is governed by yield and fracture of the anchor steel or by 
breakout of a roughly conical volume of masonry starting at the anchor head and having a fracture 
surface oriented at 45 degrees to the masonry surface. Steel capacity is calculated using the effective 
tensile stress area of the anchor (that is, including the reduction in area of the anchor shank due to 
threads). Masonry breakout capacity is calculated using expressions adapted from concrete design, 
which use a simplified design model based on a stress of 4 fm' uniformly distributed over the area of 
that right circular cone, projected onto the surface of the masonry. Reductions in breakout capacity due 
to nearby edges or adjacent anchors are computed in terms of reductions in those projected areas 
(Brown and Whitlock, 1983). 
The tensile capacity of a bent-bar anchor bolt (J- or L-bolt) is governed by yield and fracture of the 
anchor steel, by tensile cone breakout of the masonry, or by straightening and pullout of the anchor from 
the masonry. Capacities corresponding to the first two failure modes are calculated as for headed 
anchor bolts. Pullout capacity is calculated as proposed by Shaikh (1996). Possible contributions to 
tensile pullout capacity due to friction are neglected. 
 Masonry Structure Design Requirements 
225 
The tensile breakout capacity of a headed anchor is usually much greater than the pullout capacity of a 
J- or L-bolt. The designer is encouraged to use headed anchors when anchor tensile capacity is critical. 
The shear capacity of a headed or a bent-bar anchor bolt is governed by yield and fracture of the anchor 
steel or by masonry shear breakout. Steel capacity is calculated using the effective tensile stress area 
(that is, threads are conservatively assumed to lie in the critical shear plane). Shear breakout capacity is 
calculated as proposed by Brown and Whitlock, 1983. 
Under static shear loading, bent-bar anchor bolts (J- or L-bolts) do not exhibit straightening and pullout. 
Under reversed cyclic shear, however, available research suggests that straightening and pullout may 
occur. Headed anchor bolts are recommended for such applications (Malik et al., 1982). 
REFERENCES 
Abboud, B. E., X. Lu, and F. C. Schmitt. 1993. An Evaluation of Three Current Deflection Methods 
for Predicting the Lateral Deflection of Masonry Walls. In Proceedings of the Sixth North American 
Masonry Conference, Philadelphia, Pennsylvania, pp.73-85. 
Abboud, B. E., and A. A. Hamid. 1987. A Comparative Study of the Flexural Behavior of Reinforced 
Block Masonry and Reinforced Concrete Using Small Scale Model Walls. In Proceedings of the 
Fourth North American Masonry Conference, Los Angeles, California. 
Abboud, B. E., A. A. Hamid, and H. G. Harris. 1990. Small Scale Modeling of Concrete Block 
Masonry Structures, ACI Structural Journal 87(2):145-155. 
Abboud, B. E. 1987. The Use of Small Scale Direct Models for Concrete Block Masonry Assemblages 
and Slender Reinforced Walls Under Out-of-Plane Loads, Doctoral Thesis, Drexel University, 
Philadelphia, Pennsylvania. 
Blondet, J. M., R. L. Mayes, T. E. Kelley, R. R. Villablanca, and R. E. Klingner. 1989. Performance of 
Engineered Masonry in the Chilean Earthquake of March 3, 1985: Implications for U.S. Design 
Practice. Austin: Phil Ferguson Structural Engineering Laboratory, University of Texas. 
Brown, R. H., and A. R. Whitlock. 1983. Strength of Anchor Bolts in Grouted Masonry, ASCE 
Journal of Structural Engineering 109(6). 
CEB. 1995. Design of Fastenings in Concrete (Draft CEB Guide, Parts 1 to 3) and Fastenings for 
Seismic Retrofitting (State-of-the-Art Report on Design and Application), Task Group 3.5 
(Embedments), Euro-International Concrete Committee (CEB), CEB Bulletin D=Information 226. 
Comite Euro-International du Beton. 
Chen, S. W., P. A. Hidalgo, R. L. Mayes, R. W. Clough, and H. D. McNiven. 1978. Cyclic Loading 
Tests of Masonry Single Piers, Vol. 2, EERC Report UCB/EERC-78/27. Berkeley: University of 
California Earthquake Engineering Research Center. 
Fattal, S. G. 1993a. Strength of Partially Grouted Masonry Shear Walls, Report NIST 5147. 
Gaithersburg, Maryland: National Institute of Standards and Technology. 
Fattal, S. G. 1993b. The Effect of Critical Parameters on the Behavior of Partially Grouted Masonry 
Shear Walls Under Lateral Loads, NIST Report NISTIR 5116. Gaithersburg, Maryland: National 
Institute of Standards and Technology. 
Fattal, S. G., and D. R. Todd. 1991. Ultimate Strength of Masonry walls: Prediction vs. Test Results, 
NIST Report NISTIR 4633. Gaithersburg, Maryland: National Institute of Standards and Technology. 
Hamid, A. A., M. K. Hatem, H. G. Harris, and B. E. Abboud. 1990. Hysteretic Response and 
Ductility of Reinforced Concrete Masonry Walls Under Out-of-Plane Loading. In Proceedings of the 
Fifth North American Conference, Urbana, Illinois, pp. 397-410. 
2 003 Commentary, Chapter 11 
226 
Hamid, A. A., B. E. Abboud, M. W. Farah, M. K. Hatem, and H. G. Harris. 1989. Response of 
Reinforced Block Masonry Walls to Out-of-Plane Static Loads. In U.S.-Japan Coordinated Program 
For Masonry Building Research, Report 3.2(a). Drexel University. 
He, L., and N. Priestley. 1992. Seismic Behavior of Flanged Masonry Shear Walls - Final Report, 
TCCMaR Report 4.1-2. 
Hidalgo, P. A., R. L. Mayes, H. D. McNiven, and R. W. Clough. 1978. Cyclic Loading Tests of 
Masonry Single Piers, Vol. 1 and 3, EERC Report UCB/EERC-78/27 and 79/12. Berkeley, University 
of California Earthquake Engineering Research Center. 
Horton, R. T., and M. K. Tadros. 1990. Deflection of Reinforced Masonry Members, ACI Structural 
Journal 87(4):73-85. 
Jail, I., W. Kelm, and R. E. Klingner. 1992. Performance of Masonry and Masonry Veneer Buildings 
in the Loma Prieta Earthquake, PMSEL Report 92-1. Austin: University of Texas, Department of 
Civil Engineering. 
Klingner, R. E. 1994. Performance of Masonry Structures in the Northridge, California, Earthquake of 
January 17, 1991. Boulder, Colorado: The Masonry Society. 
Leiva, G., and R. Klingner. 1991. In-Plane Seismic Resistance of Two-story Concrete Masonry Shear 
Walls with Openings, TCCMaR Report 3.1(c)-2. 
Limin, H., and N. Priestley. 1988. Seismic Behavior of Flanged Masonry Shear Walls, TCCMaR 
Report 4.1-1. 
Malik, J. B., J. A. Mendonca, and R. E. Klingner. 1982. Effect of Reinforcing Details on the Shear 
Resistance of Short Anchor Bolts Under Reversed Cyclic Loading, Journal of the American Concrete 
Institute, Proceedings, 79, (1): 3-11. 
Matsumura, A. 1987. Shear Strength of Reinforced Hollow Unit Masonry Walls. In Proceedings of 
the 4th North American Masonry Conference, Los Angeles, California. 
Mayes, R. L., Y. Omote, and R. W. Clough. 1976. Cyclic Shear Tests of Masonry Piers, Vol. 1, Report 
76-8. Berkeley: University of California Earthquake Engineering Research Center. 
Merryman, K., G. Leiva, N. Antrobus, and R. Klingner. 1989. In-Plane Seismic Resistance of Twostory 
Concrete Masonry Coupled Shear Walls, TCCMaR Report 3.1(c)-1. 
Okamoto, S., Y. Yamazaki, T. Kaminosono, M. Teshigawara, and H. Hirashi. 1987. Seismic Capacity 
of Reinforced Masonry Walls and Beams in Wind and Seismic Effects. In Proceedings of the 18th 
joint Meeting, U.S. - Japan Panel on Wind and Seismic Effects, ed. by N. J. Raufaste, Report NBSIR 
87-3540. Gaithersburg, Maryland: National Institute of Standards and Technology. 
Priestley, N., and H. Limin. 1990. Seismic Response of T-Section Masonry Shear Walls. TMS 
Journal 9(1):10-19. 
Seible, F., G. A. Hegemier, M. J. N. Priestley, G. R. Kingsley, A. Kurkchubasche, and A. Igarashi. 
1992. The U.S. - TCCMaR Five-story Full Scale Masonry Research Building Test - Preliminary Report, 
TCCMaR Report 9.2-4. 
Shaikh, A. Fattah. 1996. Design of Hooked Anchor bolts in Concrete and Masonry: Proposed Code 
Provisions and Commentary. Prepared for the National Codes and Standards Council, 1996. 
Shing, P. B., J. Noland, H. Spaeh, E. Klamerus, and M. Schuller. 1991. Response of Singlestory 
Reinforced Masonry Shear Walls to In-Plane Lateral Loads, TCCMaR Report 3.1(a)-2. 
Shing, P. B., M. Schuller, and V. S. Hoskere. 1990. In-Plane Resistance of Masonry Shear Walls, 
ASCE Journal of Structural Engineering 116(3). 
 Masonry Structure Design Requirements 
227 
Sveinsson, B. I., H. D. McNiven, and H. Sucuoglu. 1985. Cyclic Shear Tests of Masonry Piers, Vol. 4, 
Report UCB/EERC 85-15. Berkeley: University of California Earthquake Engineering Research 
Center. 
Wolde-Tinsae, A. M., R. H. Atkinson, and A. A. Hamid. 1993. State-of-the-Art Modulus of Elasticity 
of Masonry. In Proceedings of the Sixth North American Masonry Conference. Boulder, Colorado: 
The Masonry Society. 
The following Technical Coordinating Committee for Masonry Research task reports not specifically 
cited but providing the substantiating data for the strength design criteria presented in this chapter are 
available through the Earthquake Engineering Research Center Library in Richmond, California (phone 
415-231-9403): 
Task No. Author(s) and Title 
1.1-1: Atkinson and Kingsley, Comparison of the Behavior of Clay & Concrete Masonry in 
Compression, September 1985. 151 pgs. 
1.2(a)-1: Hamid, A. A., G. F. Assis, and H. G. Harris, Material Models for Grouted Block Masonry, 
August 1988. 67 pgs. 
1.2(a)-2: Assis, G. F., A. A. Hamid, and H. G. Harris, Material Models for Grouted Block Masonry, 
August 1989. 134 pgs. 
1.2(b)-1: Young, J. M., and R. H. Brown, Compressive Stress Distribution of Grouted Hollow Clay 
Masonry Under Strain Gradient, May 1988, 170 pgs. 
1.3-1: Atkinson, R. H., An Assessment of Current Material Test Standards for Masonry Limit 
States Design Methods, June 1991. 38 pgs. 
2.1-1: Hart, G., and M. Basharkhah, Slender Wall Structural Engineering Analysis Computer 
Program (Shwall, Version 1.01), September 1987. 68 pgs. 
2.1-2: Hart, G., and M. Basharkhah, Shear Wall Structural Engineering Analysis Computer 
Program (Shwall, Version 1.01). September 1987. 75 pgs. 
2.1-3: Nakaki, D., and G. Hart, Uplifting Response of Structures Subjected to Earthquake Motions, 
August 1987. 200 pgs. 
2.1-4: Hart, G., N. Sajjad, and M. Basharkhah, Inelastic Column Analysis Computer Program 
(INCAP, Version 1.01), March 1988. 
2.1-5: Hong, W. K., G. C. Hart, and R. E. Englekirk, Force-Deflection Evaluation and Models for 
University of Colorado Flexural Walls, December 1989. 
2.1-6: Hart, G. C., J. W. Jaw, and Y. K. Low, SCM Model for University of Colorado Flexural 
Walls, December 1989. 31 pgs. 
2.1-7: Hart, G. C., N. Sajjad, and M. Basharkhah, Inelastic Masonry Flexural Shear Wall Analysis 
Computer Program, February 1990. 41 pgs. 
2.1-8: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic Performance 
Study, DPC Gymnasium, Elastic Time History Analysis Using SAP90, February 1992. 41 
pgs. 
2.1-9: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic Performance 
Study, TMS Shopping Center, Elastic Time History Analysis Using SAP90, February 1992. 
42 pgs. 
2.1-10: Hart, G. C., R. Englekirk, J. W. Jaw, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic 
Performance Study, RCJ Hotel, February 1992. 51 pgs. 
2 003 Commentary, Chapter 11 
228 
2.1-11: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Performance Study, 
2-Story Masonry Wall-Frame Building, February 1992. 112 pgs. 
2.1-12: Hart, G. C., R. Englekirk, J. W. Jaw, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic 
Performance Study, Designed by Tentative Limit Sates Design Standard, February 1992. 
75 pgs. 
2.2-1: Ewing, R. D., A. El-Mustapha, and J. Kariotis, FEM/I - A Finite Element Computer 
Program for the Nonlinear Static Analysis of Reinforced Masonry Building Components, 
December 1987 (Revised June 1990). 124 pgs. 
2.2-2: Ewing, R. D., Parametric Studies on Reinforced Masonry Shear Walls Using FEM/I, A 
Nonlinear Finite Element Analysis Program, March 1992. 
2.2-3: Ewing, R. D., Finite Element Analysis of Reinforced Masonry Building Components 
Designed by a Tentative Masonry Limit States Design Standard, March 1992. 48 pgs. 
2.3-1: Ewing, R., J. Kariotis, and A. El-Mustapha, LPM/I, A Computer Program for the 
Nonlinear, Dynamic Analysis of Lumped Parameter Models, August, 1987. 200 pgs. 
2.3-2: Kariotis, J., A. El-Mustapha, and R. Ewing, Influence of Foundation Model on the Uplifting 
of Structures, July 1988. 50 pgs. 
2.3-3: Kariotis, J., A. Rahman, and A. El-Mustapha, Investigation of Current Seismic Design 
Provisions for Reinforced Masonry Shear Walls, January 1990. 48 pgs. 
2.3-4: Kariotis, J., A. Rahman, O. Waqfi, and R. Ewing, Version 1.03 LPM/I - A Computer 
Program for the Nonlinear, Dynamic Analysis of Lumped Parameter Models, February 
1992. 227 pgs. 
2.3-5: Kariotis, J., O. Waqfi, and R. Ewing, R., A Computer Program Using Beam Elements for 
the Nonlinear, Dynamic Analysis of Lumped Parameter Models, February 1992. 96 pgs. 
2.3-6: Kariotis, J., and O. Waqfi, Comparison of the Dynamic Response of a Damped MDOF 
Nonlinear Beam Model with an Equivalent SDOF Hysteretic Model, April 1992. 88 pgs. 
2.3-7: Kariotis, J., and O. Waqfi, Recommended Procedure for Calculation of the Balanced 
Reinforcement Ratio, February 1992. 73 pgs. 
2.4(b)-1: Button, M. R., and R. L. Mayes, Out-of-Plane Seismic Response of Reinforced Masonry 
Walls: Correlation of Full-Scale Test and Analytical Model Results, March 1991. 65 pgs. 
3.1(a)-1: Scrivener, J., Summary of Findings of Cyclic Tests on Masonry Piers, June 1986. 7 pgs. 
3.1(a)-2: Shing, P. B., J. Noland, H. Spaeh, E. Klamerus, and M. Schuller, Response of Single-Story 
Reinforced Masonry Shear Walls to In-Plane Lateral Loads, January 1991. 136 pgs. 
3.1(b)-1: Seible, F., and H. LaRovere, Summary of Pseudo Dynamic Testing, February 1987. 46 pgs. 
3.1(b)-2: Igarashi, A., F. Seible, and G. Hegemier, Development of the Generated Sequential 
Displacement Procedure and the Simulated Seismic Testing of the TCCMaR Three-Story In- 
Plane Walls, June 1993. 
3.1(c)-1: Merryman, K., G. Leiva, B. Antrobus, and R. Klingner, In-Plane Seismic Resistance of 
Two-Story Concrete Masonry Coupled Shear Walls, September 1989. 176 pgs. 
3.1(c)-2: Leiva, G., and R. Klingner, In-plane Seismic Resistance of Two-story Concrete Masonry 
Shear Walls with Openings, August 1991. 326 pgs. 
3.2(a)-1: Hamid, A., B. Abboud, M. Farah, K. Hatem, and H. Harris, Response of Reinforced Block 
Masonry Walls to Out-of-Plane Static Loads, September 1989. 120 pgs. 
 Masonry Structure Design Requirements 
229 
3.2(b)-1: Agbabian, M., S. Adham, S. Masri, V. Avanessian, and V. Traina, Out-of-Plane Dynamic 
Testing of Concrete Masonry Walls, Vol. 1 and 2, July 1989. 220 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 1: General Information, April 1991. 172 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 2: Walls No. 4 and 6 (Group 1), April 1991. 267 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 3: Walls No. 8, 9, 10 and 11 (Group 2), April 1991. 
310 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 4: Walls No. 3, 5, and 7 (Group 3), April 1991. 256 
pgs. 
4.1-1: He, L., and M. J. N. Priestley, Seismic Behavior of Flanged Masonry Shear Walls, May 
1988. 119 pgs. 
4.1-2: He, L., and M. J. N. Priestley, Seismic Behavior of Flanged Masonry Shear Walls - Final 
Report, November 1992. 279 pgs. 
4.2-1: Hegemier, G., and H. Murakami, On the Behavior of Floor-to-Wall Intersections in 
Concrete Masonry Construction, Part I: Experimental. 
4.2-2: Hegemier, G., and H. Murakami, On the Behavior of Floor-to-Wall Intersections in 
Concrete Masonry Construction, Part II: Theoretical. 
5.1-1: Porter, M., and A. Sabri, Plank Diaphragm Characteristics, July 1990. 226 pgs. 
5.2-1: Porter, M., F. Yeomans, and A. Johns, Assembly of Existing Diaphragm Data, July 1990. 
142 pgs. 
6.2-1: Scrivener, J., Bond of Reinforcement in Grouted Hollow-Unit Masonry: A State-of-the-Art, 
June 1986. 53 pgs. 
6.2-2: Soric, Z., and L. Tulin, Bond Splices in Reinforced Masonry, August 1987. 296 pgs. 
7.1-1: Paulson, T., and D. Abrams, Measured Inelastic Response of Reinforced Masonry Building 
structures to Earthquake Motions, October 1990. 294 pgs. 
8.1-1: Hart, G., A Limit State Design Method for Reinforced Masonry, June 1988. 
8.1-2: Hart, G., Expected Value Design in the Context of a Limit Sate Design Methodology, 
February 1990. 
8.2-1: Hart, G., and G. T. Zorapapel, Reliability of Concrete Masonry Wall Structures, December 
1991. 229 pgs. 
8.2-2: Hart, G., and N. Sajjad, Confinement in Concrete Masonry, December 1990. 
8.2-3: Hart, G., and J. Jang, Seismic Performance of Masonry Wall Frames, December 1991. 
9.1-1: Kariotis, J. C., and A. W. Johnson, Design of Reinforced Masonry Research Building, 
September 1987. 42 pgs. 
9.1-2: Kariotis, J. C., and O. M. Waqfi, Trial Designs Made in Accordance with Tentative Limit 
States Design Standards for Reinforced Masonry Buildings, February 1992. 184 pgs. 
9.2-1: Seible, F., Report on Large Structures Testing Facilities in Japan, September 1985. 120 
pgs. 
2 003 Commentary, Chapter 11 
230 
9.2-2: Seible, F., Design and Construction of the Charles Lee Powell Structural Systems 
Laboratory, November 1986. 65 pgs. 
9.2-3: Seible, F., The Japanese Five-story Full Scale Reinforced Masonry Building Test, January 
1988. 100 pgs. 
9.2-4: Seible, F., G. A. Hegemier, M. J. N. Priestley, G. R. Kingsley, A. Kurkchubasche, and A. 
Igarashi, The U.S. - TCCMaR Five-story Full Scale Masonry Research Building Test - 
Preliminary Report, October 1992. 58 pgs. 
11.1-1: TCCMaR, Summary Report: U.S. Coordinated Program for Masonry Building Research, 
September 1985 to August 1986. 190 pgs. 
11.1-2: TCCMaR, Status Report: U.S. Coordinated Program for Masonry Building Research, 
November 1988. 170 pgs. 

 231 
Chapter 12 Commentary 
WOOD STRUCTURE DESIGN REQUIREMENTS 
12.1 GENERAL 
12.1.2 References. Wood construction practices have not been codified in a form that is standard 
throughout the country. The 2003 Provisions incorporates by reference the AF&PA ASD/LRFD 
Supplement, Special Design Provisions for Wind and Seismic (SDPWS) and the 2003 International 
Residental Code (IRC). Many wood frame structures are a combination of engineered wood and 
conventional light-frame construction. Wood also is used in combination with other materials 
(American Institute of Timber Construction, 1985; Breyer, 1993; Faherty and Williamson, 1989; 
Hoyle and Woeste, 1989; Somayaji, 1992; Stalnaker and Harris, 1989). The requirements of the 
model building codes were used as a resource in developing the requirements introduced in the 1991 
Provisions and further modified since then. The general requirements of Chapter 12 cover construction 
practices necessary to provide a performance level of seismic resistance consistent with the purposes 
stated in Chapter 1. These requirements also may be related to gravity load capacity and wind force 
resistance which is a natural outgrowth of any design procedure. For the 2003 Provisions, the 
reference documents continue to be grouped according to their primary focus into three subsections: 
Sec. 12.1.2.1, Engineered Wood Construction; Sec. 12.1.2.2, Conventional Construction; and Sec. 
12.1.2.3, Materials Standards. 
12.2 DESIGN METHODS 
Prior to the publication of AF&PA/ASCE 16, typical design of wood frame structures followed the 
American Forest and Paper Association (AF&PA) National Design Specification for Wood 
Construction (NDS) (AF&PA, 1991). The NDS is based on allowable stresses and implied factors 
of safety. However, the design procedure provided by the Provisions was developed on the premise of 
the resistance capacity of members and connections at the yield level (ASCE, 1988; Canadian Wood 
Council, 1990 and 1991; Keenan, 1986). In order to accommodate this difference in philosophy, the 
1994 and prior editions of the Provisions made adjustments to the tabulated allowable stresses in the 
reference documents. 
With the completion of the Load and Resistance Factor Standard for Engineered Wood Construction 
(AF&PA/ASCE, 1995), the modifications and use of an allowable stress based standard was no 
longer necessary. Therefore, the 1997 Provisions included the LRFD standard by reference 
(AF&PA/ASCE 16) and used it as the primary design procedure for engineered wood construction. 
The use of AF&PA/ASCE 16 continues in the 2003 Provisions. 
Conventional light-frame construction, a prescriptive method of constructing wood structures, is 
allowed for some design categories. These structures must be constructed according to the 
requirements set forth in Sec. 12.4 and applicable reference documents. If the construction deviates 
from these prescriptive requirements, the engineered design requirements of Sec. 12.2 and 12.3 and 
AF&PA/ASCE 16 must be followed. If a structure that is classified as conventional construction 
contains some structural elements that do not meet the requirements of conventional construction, the 
elements in question can be engineered without changing the rest of the structure to engineered 
construction. The extent of design to be provided must be determined by the responsible registered 
design professional; however, the minimum acceptable extent is often taken to be force transfer into 
the element, design of the element, and force transfer out of the element. This does not apply to a 
structure that is principally an engineered structure with minor elements that could be considered 
conventional. When more than one braced wall line or diaphragm in any area of a conventional 
residence requires design, the nature of the construction may have changed, and engineered design 
2003 Commentary, Chapter 12 
232 
might be appropriate for the entire seismic-force-resisting system. The absence of a ceiling diaphragm 
may also create a configuration that is non-conventional. The requirement for engineering portions of 
a conventional construction structure to maintain lateral-force resistance and stiffness is added to 
provide displacement compatibility. 
Alternate strength of members and connections. It remains the intent of the Provisions that load 
and resistance factor design be used. When allowable stress design is to be used, however, the 
factored resistance of members and connections subjected to seismic forces acting alone or in 
combination with other prescribed loads shall be determined using a capacity reduction factor, f, times 
2.16 times the allowable stresses permitted in the National Design Specification for Wood 
Construction (NDS) and supplements (AF&PA, 1991). The allowable stresses used shall not include a 
duration of load factor, CD. The value of the capacity reduction factor, f, shall be as follows: 
Wood members 
In flexure f = 1.00 
In compression f = 0.90 
In tension f = 1.00 
In shear and torsion f = 1.00 
Connectors 
Anchor bolts, bolts, lag bolts, nails, screws, etc. f = 0.85 
Bolts in single shear in members of a 
seismic-force-resisting system f = 0.40 
These soft conversions from allowable stress design values to load and resistance factor design 
values first appeared in Sec. 9.2 in the 1994 Provisions. An alternative method of calculating soft 
conversions is provided in ASTM D 5457-93. The reader is cautioned, however, that the loads and 
load combinations to be used for conversion are not specified so it is incumbent upon the user to 
determine appropriate conversion values. Wood frame structures assigned to Seismic Design 
Category A, other than one- and two-family dwellings, must comply with Sec. 12.4 or if engineered 
need only comply with the reference documents and Sec. 1.5. Exceptions addressing one- and twofamily 
detached dwellings appear in Sec. 
12.2.1 Seismic Design Categories B, C, and D. Seismic Design Categories B, C, and D were 
combined in the 1997 Provisions. At the same time, subsections on material limitations and 
anchorage requirements were moved. This was based on the philosophy that detailing requirements 
should vary based on R value rather than seismic design category. 
Structures assigned to Seismic Design Categories B, C, and D are required to meet the minimum 
construction requirements of Sec. 12.4 (Sherwood and Stroh, 1989) or must be engineered using 
standard design methods and principles of mechanics. Conventional light-frame construction 
requirements were modified in the 1991 Provisions to limit the spacing between braced wall lines 
based on calculated capacities to resist the loads and forces imposed. 
Engineered structures assigned to Seismic Design Categories B, C, and D are required to conform to 
the provisions of Sec. 12.2 and 12.3. Included in these sections are general design limitations, limits 
on wood resisting forces contributed by concrete or masonry, shear wall and diaphragm aspect ratio 
limitations, and requirements for distribution of shear to vertical resisting elements. 
Wood Structure Design Requirements 
233 
12.2.2 Seismic Design Categories E and F. If the provisions of Chapter 12 apply, Seismic Design 
Category E and F structures require an engineered design. Conventional construction is not 
considered rigorous enough for structures expected to be functional following a major seismic event. 
For Seismic Design Category E and F structures, close attention to load path and detailing is required. 
Structures assigned to Seismic Design Category E and F require blocked diaphragms. Structural-use 
panels must be applied directly to the framing members; the use of gypsum wallboard between the 
structural-use panels and the framing members is prohibited because of the poor performance of nails 
in gypsum. Restrictions on allowable shear values for structural-use shear panels when used in 
conjunction with concrete and masonry walls are intended to provide for deformation compatibility of 
the different materials. 
12.2.3.1 Discussion of cyclic test protocol is included in ATC (1995), Dolan (1996), and Rose (1996). 
12.2.3.2 and 12.2.3.7 The mid-span deflection of a simple-span, blocked wood structural panel 
diaphragm uniformly nailed throughout may be calculated by use of the following formula: 
5 3 0.188 ( ) 
8 4 2 c 
n 
vL vL Le X 
bEA Gt b 
S . 
. = + + + 
where: 
. = the calculated deflection, in. (mm). 
v = maximum shear due to factored design loads in the direction under consideration, lb/ft 
(kN/m). 
L = diaphragm length, ft (m). 
b = diaphragm width, ft (m). 
E = elastic modulus of chords, psi (MPa). 
A = area of chord cross-section, in.2 (mm2). 
Gt = panel rigidity through the thickness, lb/in. (N/mm). 
en = nail deformation, in. (mm). 
S (.cX) = sum of individual chord-splice slip values on both sides of the diaphragm, each 
multiplied by its distance to the nearest support, in. (mm). 
If not uniformly nailed, the constant 0.188 in the third term must be modified accordingly. See ATC 7 
(Applied Technology Council, 1981). 
This formula was developed based on engineering principles and monotonic testing. Therefore, it 
provides an estimate of diaphragm deflection due to loads applied in the factored resistance shear 
range. The effects of cyclic loading and resulting energy dissipation may alter the values for nail 
deformation in the third term, as well as chord splice effects of the fourth term, if mechanically-spliced 
wood chords are used. The formula is not applicable to partially-blocked diaphragms. 
The deflection of a blocked wood structural panel shear wall may be calculated by use of the following 
formula. 
8 3 0.75 n a 
vh vh he h d 
bEA Gt b 
. = + + + 
where: 
. = the calculated deflection, in. (mm). 
v = maximum shear due to factored design loads at the top of the wall, lb/ft (kN/m). 
2003 Commentary, Chapter 12 
234 
h = shear wall height, ft (m). 
b = shear wall width, ft (m). 
E = elastic modulus of boundary element (vertical member at shear wall boundary), psi (MPa). 
A = area of boundary element cross-section (vertical member at shear wall boundary), in.2 
(mm2). 
Gt = panel rigidity through the thickness, lb/in. (N/mm). 
en = nail deformation, in. (mm). 
da = deflection due to anchorage details ( rotation and slip at hold downs), in. (mm). 
Guidance for use of the above two equations can be found in the references. 
One stipulation is that there are no accepted rational methods for calculating deflections for 
diaphragms and shear walls that are sheathed with materials other than wood structural panel products 
fastened with nails. Therefore, if a rational method is to be used, the capacity of the fastener in the 
sheathing material must be validated by acceptable test procedures employing cyclic forces or 
displacements. Validation must include correlation between the overall stiffness and capacity 
predicted by principles of mechanics and that observed from test results. A diaphragm or shear wall 
sheathed with dissimilar materials on the two faces should be designed as a single-sided wall using the 
capacity of the stronger of the materials and ignoring the weaker of the materials. 
TABLE C12.2A 
en FASTENER SLIP EQUATIONS FOR USE IN CALCULATING DIAPHRAGM 
AND SHEAR WALL DEFLECTION DUE TO FASTENER SLIP 
Fastener Slip, en (in.)1 
Fastener 
Minimum 
Penetration 
(in.) 
Maximum Fastener 
Loads - Vn 
(lb/fastener) 
Fabricated w/green 
(>19% m.c.) 
lumber 
Fabricated 
w/dry (< 19% 
m.c.) lumber 
6d common nail 1-1/4 180 (Vn/434)2.314 (Vn/456)3.144 
8d common nail 1-3/8 220 (Vn/857)1.869 (Vn/616)3.018 
10d common nail 1-1/2 260 (Vn/977)1.894 (Vn/769)3.276 
14-ga staple 1 to <2 140 (Vn/902)1.464 (Vn/596)1.999 
14-ga staple >2 170 (Vn/674)1.873 (Vn/461)2.776 
For SI: 1 inch = 25.4 mm, 1 pound = 4.448 N. 
1.Values apply to plywood and OSB fastened to lumber with a specific gravity of 0.50 or greater except that the 
slip shall be increased by 20 percent when plywood is not Structural I. 

Wood Structure Design Requirements 
235 
TABLE C12.2B 
VALUES OF Gt FOR USE IN CALCULATING DEFLECTION OF 
WOOD STRUCTURAL PANEL DIAPHRAGMS AND SHEAR WALLS 

2 003 Commentary, Chapter 12 
236 
VALUES OF Gt (lb/in. panel depth or width) 
STRUCTURAL I OTHER 
PANEL 
TYPE 
Thickness 
(in.) All Plywood Grades Marine All Other 
Plywood 
Grades 
 31,000 31,000 24,000 
11/32 33,000 33,000 25,500 
3/8 34,000 34,000 26,000 
15/32 49,500 49,500 38,000 
 50,000 50,000 38,500 
19/32 63,500 63,500 49,000 
5/8 64,500 64,500 49,500 
23/32 65,500 65,500 50,500 
 66,500 66,500 51,000 
7/8 68,500 68,500 52,500 
1 95,500 95,500 73,500 
Sanded 
Plywood 
1-1/8 97,500 97,500 75,000 
For SI: 1 inch = 25.4 mm, 1 pound/inch of panel depth or width = 0.1751 N/mm. 
1. Applies to plywood with 5 or more layers; for 5 ply/3 layer plywood, use values for 4 ply. 
Effect of Green Lumber Framing on Diaphragms and Shear Walls: A recent study of wood 
structural panel shear walls (APA Report T2002-53) fabricated with wet lumber and tested when dry 
shows that shear stiffness is affected to a much larger degree than shear strength when compared to 
control specimens fabricated with dry lumber and tested when dry. The shear strength of walls 
fabricated with wet lumber showed negligible reductions (0-7 percent) when compared to control 
specimens. The shear stiffness of walls fabricated with wet lumber was always reduced when 
compared to control specimens. Observed reductions in stiffness were consistent with predicted 
stiffness reductions based on use of Eq. C12.2A and nail slip values specified in Table C12.2A. For 
example, measured deflection of a standard wall configuration at the shear wall factored unit shear 
value was approximately 2.5 times the deflection of the control specimen and predicted deflections 
were within 0.05 inches of the test deflection for both the fabricated wet specimen and control 
specimen. 
As a result of these tests, direct consideration of shear wall stiffness is recommended in lieu of 
applying shear wall strength reductions when wood structural panel shear walls are fabricated with 
wet lumber (e.g. moisture content > 19 percent). To address reduced shear stiffness for shear walls 
fabricated with wet lumber, story drift calculations should be based on en values for lumber with 
moisture content > 19 percent to determine compliance with allowable story drift limits of the 
Provisions. A similar relationship can be expected when analyzing the deflection of diaphragms. 
The designer should keep in mind that deflection equations are verified for walls with wood 
structural panel sheathing only and does not address the increased stiffness provided by finish 
materials such as gypsum and stucco. The CUREE-Caltech Woodframe project illustrated that 
finishes such as gypsum wallboard and stucco increase the stiffness of the walls. While these 
Wood Structure Design Requirements 
237 
deflection equations are currently the best estimate of wood structural panel wall deflection, actual 
wall deflections will likely be less than predicted deflections due to the presence of finish materials 
in typical wall construction. 
12.2.3.11 and 12.2.3.12. Tie-down devices should be based on cyclic tests of the connection to 
provide displacement capacity that allows rotation of the end post without significant reduction in 
the shear wall resistance. The tie-down device should be stronger than the lateral capacity of the 
wall so that the mechanism of failure is the sheathing fasteners and not a relatively brittle failure of 
the wall anchorage. For devices for which the published resistance is in allowable stress design 
values, the nominal strength shall be determined by multiplying the allowable design load by 1.3. 
The nominal strength of a tie-down device may be determined as the average maximum test load 
resisted without failing under cyclic loading. In that case, the average should be based on tests of at 
least three specimens. 
Calculations of deflection of shear walls should include the effects of crushing under the 
compression chord, uplift of the tension chord, slip in the tie-down anchor with respect to the post, 
and shrinkage effects of the platforms, which primarily consist of floor framing members. 
Movement associated with these variables can be significant and neglecting their contribution to the 
lateral displacement of the wall will results in a significant under-estimation of the deflection. 
Custom tie-down devices are permitted to be designed using methods for the particular materials 
used and AF&PA/ASCE 16 under alternative means and methods. 
Tie-down devices that permit significant vertical movement between the tie-down and the tie-down 
post can cause failure in the nails connecting the shear wall sheathing to the sill plate. High tension 
and tie-down rotation due to eccentricity can cause the bolts connecting the tie-down bracket to the 
tie-down post to pull through and split the tie-down post. Devices that permit such movement 
include heavily loaded, one-sided, bolted connections with small dimensions between elements 
resisting rotation due to eccentricity. Any device that uses over-drilled holes, such as most bolted 
connections, will also allow significant slip to occur between the device and the tie-down post 
before load is restrained. Both the NDS and the steel manual specify that bolt holes will be overdrilled 
as much as 1/16 in. (2 mm). This slip is what causes much of the damage to the nails 
connecting the sheathing to the sill plate. Friction between the tie-down post and the device cannot 
be counted on to resist load because relaxation in the wood will cause a loss of clamping and, 
therefore, a loss in friction over time. This is why all tests should be conducted with the bolts 
finger tight as opposed to tightening with a wrench. 
Cyclic tests of tie-down connections must follow a pattern similar to the sequential phased 
displacement (SPD) tests used by Dolan (1996) and Rose (1996). These tests used full wall 
assemblies and therefore induced deflection patterns similar to those expected during an earthquake. 
If full wall assembly tests are not used to test the tie-down devices, it must be shown that the 
expected rotation as well as tension and compression are used. This is to ensure that walls using the 
devices will be able to deform in the intended manner. This allows the registered design 
professional to consider compatibility of deformations when designing the structure. 
Splitting of the bottom plate of the shear walls has been observed in tests as well as in structures 
subjected to earthquakes. Splitting of plates remote from the end of the shear wall can be caused by 
the rotation of individual sheathing panels inducing upward forces in the nails at one end of the 
panel and downward forces at the other. With the upward forces on the nails and a significant 
distance perpendicular to the wall to the downward force produced by the anchor bolt, high crossgrain 
bending stresses occur. Splitting can be reduced or eliminated by use of large plate washers 
that are sufficiently stiff to reduce the eccentricity and by use of thicker sill plates. Thicker sill 
plates (3 in. nominal, 65 mm) are recommended for all shear walls for which Table 12.2-3a (or 12.2- 
3b) requires 3 in. nominal (65 mm) framing to prevent splitting due to close nail spacing. This is to 
help prevent failure of the sill plate due to high lateral loading and cross-grain bending. 
2 003 Commentary, Chapter 12 
238 
The tendency for the nut on a tie-down bracket anchor bolt to loosen significantly during cycled 
loading has been observed in some testing. One tested method of limiting the loosening is to apply 
adhesive between the nut and tie-down bolt. 
A logical load path for the structure must be provided so that the forces induced in the upper 
portions of the structure are transmitted adequately through the lower portions of the structure to the 
foundation. 
In the 2003 Provisions update cycle anchorage provisions were divided into two distinct subsections 
to separately address anchorage for uplift and anchorage for in-plane shear. The title section was 
clarified to address both traditional segmented shear walls and perforated shear walls. 
A prior Provisions requirement that nuts on both uplift anchors and in-plane shear anchors be 
prevented from loosening prior to covering the framing, was deleted. This provision was originally 
based on observed backing-off of nuts in a small number of cyclic tests of shear walls but in the 
large number of tests conducted since that time this phenomenon has not been observed to occur. It 
was felt that retaining the existing requirement for tightening the nuts prior to closing in the framing 
was sufficient to address this issue. 
A prior Provisions requirement for the nominal strength of a tiedown to be equal to or exceed the 
factored resistance of the shear wall times Oo 
/ 1.3, was replaced with simpler wording that has an 
equivalent effect and is intended primarily as a statement of design philosophy. The new language 
in Sec 12.2.3.11 only refers to the nominal strength of the tiedown and the nominal strength of the 
shear wall. Nominal strengths for typical nailed wood structural panel shear walls are set forth in 
Table 4.3A column B of AF&PA ASD/LRFD Supplement, Special Design Provisions for Wind and 
Seismic. In addition, similar language making the nominal strength of in-plane shear anchorage 
match the nominal strength values of the shear walls was added, to provide a basis for design of inplane 
shear connections that is consistent with requirements for uplift anchorage. The capacity-based 
nominal strength have been introduced primarily as a statement of design philosophy, with the intent 
of forcing sheathing nailing to be the controlling failure mechanism. The complexity of load paths in 
wood frame buildings suggest that additional study is needed to achieve reliable development of 
desired failure mechanisms. 
Plate washers are now specifically permitted to have a diagonal slot not exceeding 1-3/4 inches in 
length to facilitate placement within the width of the sill plate. 
12.2.3.14 Sheathing nails should be driven flush with the surface of the panel, and not further. This 
could result in the nail head creating a small depression in, but not fracturing, the first veneer. This 
requirement is imposed because of the significant reduction in capacity and ductility observed in 
shear walls constructed with over-driven nails. It is advised that the edge distance for sheathing 
nails be increased as much as possible along the bottom of the panel to reduce the potential for the 
nails to pull through the sheathing. 
12.3 GENERAL DESIGN REQUIREMENTS FOR ENGINEERED WOOD 
CONSTRUCTION 
Engineered construction for wood structures as defined by the Provisions encompasses all structures 
that cannot be classified as conventional construction. Therefore, any structure exceeding the height 
limitations or having braced walls spaced at intervals greater than those prescribed in Table 12.4-1 
or not conforming to the requirements in Sec. 12.4 must be engineered using standard design 
methods and principles of mechanics. Framing members in engineered wood construction are sized 
based on calculated capacities to resist the loads and forces imposed. Construction techniques that 
utilize wood for lateral force resistance in the form of diaphragms or shear walls are discussed 
further in Sec. 12.4. Limitations have been set on the use of wood diaphragms that are used in 
combination with concrete and masonry walls or where torsion is induced by the arrangement of the 
vertical resisting elements. A load path must be provided to transmit the lateral forces from the 
diaphragm through the vertical resisting elements to the foundation. It is important for the registered 
Wood Structure Design Requirements 
239 
design professional to follow the forces down, as for gravity loads, designing each connection and 
member along the load path. 
Although wood moment resisting frames are not specifically covered in the Provisions, they are not 
excluded by them. There are several technical references for their design, and they have been used 
in Canada, Europe, and New Zealand. Wood moment resisting frames are designed to resist both 
vertical loads and lateral forces. Detailing at columns to beam/girder connections is critical in 
developing frame action and must incorporate effects of member shrinkage. Detailed information 
can be obtained from the national wood research laboratories. There are many references that 
describe the engineering practices and procedures used to design wood structures that will perform 
adequately when subjected to lateral forces. The list at the end of this Commentary chapter gives 
some, but by no means all, of these. 
Deformation compatibility The registered design professional should visualize the deformed shape 
of the structure to ensure that the connections provide the necessary ductility to allow the probable 
deflection demand placed on the structure. Unlike steel or other metal structures, wood is not a 
ductile material and virtually all of the ductility achieved in the structure is in the connections. The 
planned failure mechanism of wood structures must be through the connections, including the 
nailing of structural panels; otherwise the failure will be brittle in nature. The philosophy of strong, 
elastic columns and yielding beams cannot be projected from steel to wood structures. To enable a 
wood structure to deform and dissipate energy during a seismic event, the connections must be the 
weak link in the structure and must be ductile. Recent earthquakes, such as that in Northridge, 
California, have shown failures due to the fact that consideration of deformation compatibility was 
neglected. 
As an example of a compatibility issue, consider the deformation compatibility between a tie-down 
connector to the tie-down post and the edge nailing of shear wall sheathing to the tie-down post and 
adjacent bottom plate. Recent testing and observations from the Northridge earthquake have 
suggested that the tie-down post experiences notable displacement before significant load can be 
carried through the tie-down connector. This is due, among other things, to the oversizing of the 
bolt holes in the tie-down post and the deformation and rotation of the tie-down bracket. Anchor 
bolts connecting the bottom plate to the foundation below tend to attempt to carry the shear wall 
uplift as the tie-down post moves. The sheathing, however, is nailed to both the bottom plate, which 
is held in place, and the tie-down post, which is being pulled up. The result is a large deformation 
demand being placed on the nails connecting the sheathing to the framing. This often results in the 
nails pulling out of the sheathing at the tie-down post corner and sometimes results in an unzipping 
effect where a significant portion of the remaining sheathing nailing fails as high loads cause one 
nailed connection to fail and move on to overstress the next nail. The most effective solution 
currently known is to limit the slip and deformation at the tie-down post by using a very stiff nailed 
or screwed tie-down. 
Because this is an area where understanding of compatibility issues is just starting to develop, the 
Sec. 12.3.2 provision uses the wording shall be considered in design in lieu of the originally 
proposed provision shall be made to ensure The intent is to provide guidance while not 
requiring the impossible. 
If necessary, the stiffness of the wood diaphragms and shear walls can be increased with the use of 
adhesives (if adhesives are to be used). However, it should be noted that there are no rational 
methods for determining deflections in diaphragms that are constructed with non-wood sheathing 
materials. If the nail stiffness values or shear stiffness of non-wood sheathing materials is 
determined in a scientific manner, such as through experimental cyclic testing, the calculations for 
determining the stiffness of shear panels will be considered validated. 

2 003 Commentary, Chapter 12 
240 
Limitation on forces contributed by concrete or masonry. Due to the significant difference in inplane 
stiffness between wood and masonry or concrete systems, the use of wood members to resist 
the seismic forces produced by masonry and concrete is not allowed. This is due to the probable 
torsional response such a structure will exhibit. There are two exceptions where wood can be 
considered to be part of the seismic-load-resisting system. The first is where the wood is in the form 
of a horizontal truss or diaphragm and the lateral loads do not produce rotation of the horizontal 
member. The second exception is in structures of two stories or less in height. In this case, the 
capacity of the wood shear walls will be sufficient to resist the lower magnitude loads imposed. 
Five restrictions are imposed on these structures to ensure that the structural performance will not 
include rotational response and that the drift will not cause failure of the masonry or concrete 
portions of the structure. 
Shearwalls and Diaphragms. Many wood-framed structures resist seismic forces by acting as a 
box system. The forces are transmitted through diaphragms, such as roofs and floors, to reactions 
provided by shear walls. The forces are, in turn, transmitted to the lower stories and to the final 
point of resistance, the foundations. A shear wall is a vertical diaphragm generally considered to act 
as a cantilever from the foundation. 
A diaphragm is a nearly horizontal structural unit that acts as a deep beam or girder when flexible in 
comparison to its supports and as a plate when rigid in comparison to its supports. The analogy to a 
girder is somewhat more appropriate since girders and diaphragms are made up as assemblies 
(American Plywood Association, 1991; Applied Technology Council, 1981). Sheathing acts as the 
web to resist the shear in diaphragms and is stiffened by the framing members, which also provide 
support for gravity loads. Flexure is resisted by the edge elements acting like flanges to resist 
induced tension or compression forces. The flanges may be top plates, ledgers, bond beams, or 
any other continuous element at the perimeter of the diaphragm. 
The flange (chord) can serve several functions at the same time, providing resistance to loads and 
forces from different sources. When it functions as the tension or compression flange of the 
girder, it is important that the connection to the web be designed to accomplish the shear 
transfer. Since most diaphragm flanges consist of many pieces, it is important that the splices be 
designed to transmit the tension or compression occurring at the location of the splice and to 
recognize that the direction of application of seismic forces can reverse. It should also be 
recognized that the shear walls parallel to the flanges may be acting with the flanges to distribute the 
diaphragm shears. When seismic forces are delivered at right angles to the direction considered 
previously, the flange becomes a part of the reaction system. It may function to transfer the 
diaphragm shear to the shear wall(s), either directly or as a drag strut between segments of shear 
walls that are not continuous along the length of the diaphragm. 
For shear walls, which may be considered to be deep vertical cantilever beams, the flanges are 
subjected to tension and compression while the webs resist the shear. It is important that the 
flange members, splices at intermediate floors, and the connection to the foundation be detailed 
and sized for the induced forces. 
The webs of diaphragms and shear walls often have openings. The transfer of forces around 
openings can be treated similarly to openings in the webs of steel girders. Members at the edges of 
openings have forces due to flexure and the higher web shear induced in them and the resultant 
forces must be transferred into the body of the diaphragm beyond the opening. 
In the past, wood sheathed diaphragms have been considered to be flexible by many registered 
design professionals and model code enforcement agencies. The newer versions of the model codes 
now recognize that the determination of rigidity or flexibility for determination of how forces will be 
distributed is dependent on the relative deformations of the horizontal and vertical force-resisting 
elements. Wood sheathed diaphragms in structures with wood frame shear walls with various types 
of sheathing may be relatively rigid compared with the vertical resisting system and, therefore, 
capable of transmitting torsional lateral forces. A diaphragm is considered to be flexible if its 
Wood Structure Design Requirements 
241 
deformation is two or more times that of the vertical force-resisting elements subjected to the same 
force. 
Discussions of these and other topics related to diaphragm and shear wall design, such as cyclic 
testing and pitched or notched diaphragms, may be found in the references. 
The capacity of shear walls must be determined either from tabulated values that are based on 
experimental results or from standard principles of mechanics. The tables of allowable values for 
shear walls sheathed with other than wood or wood-based structural-use panels were eliminated in 
the 1991 Provisions as a result of re-learning the lessons from past earthquakes and testing on the 
performance of structures sheathed with these materials during the Northridge earthquake. In the 
1997 Provisions values for capacity for shear walls sheathed with wood structural panels were 
reduced from monotonic test values by 10 percent to account for the reduction in capacity observed 
during cyclic tests. This decision was reviewed for the 2000 edition of the Provisions due to the 
availability of an expanded data set of test results. The reduction was removed for the 2000 
Provisions when the effect of the test loading protocol was determined to be the cause of the initial 
perceived reductions. Capacities for diaphragms were not reduced from the monotonic test values 
because the severe damage that occurred in shear walls has not been noted in diaphragms in recent 
earthquakes. 
The Provisions are based on assemblies having energy dissipation capacities which were recognized 
in setting the R factors. For diaphragms and shear walls utilizing wood framing, the energy 
dissipation is almost entirely due to nail bending. Fasteners other than nails and staples have not 
been extensively tested under cyclic load application. When screws or adhesives have been tested in 
assemblies subjected to cyclic loading, they have had a brittle mode of failure. For this reason, 
adhesives are prohibited for wood framed shear wall assemblies in SDC C and higher and only the 
tabulated values for nailed or stapled sheathing are recommended. If one wished to use shear wall 
sheathing attached with adhesives, as an alternate method of construction in accordance with Sec. 
1.1.2.5, caution should be used (Dolan and White, 1992; Foschi and Filiatrault, 1990). The 
increased stiffness will result in larger forces being attracted to the structure. The anchorage 
connections and adjoining assemblies must, therefore, be designed for these increased forces. Due 
to the brittle failure mode, these walls should be designed to remain elastic, similar to unreinforced 
masonry. The use of adhesives for attaching sheathing for diaphragms increases their stiffness, and 
could easily change the diaphragm response from flexible to rigid. 
Horizontal distribution of shear. The Provisions define when a diaphragm can be considered to 
be flexible or rigid. The purpose is to determine whether the diaphragm should have the loads 
proportioned according to tributary area or stiffness. For flexible diaphragms, the loads should be 
distributed according to tributary area whereas for rigid diaphragms, the loads should be distributed 
according to stiffness. 
The distribution of seismic forces to the vertical elements (shear walls) of the seismic-force-resisting 
system is dependent, first, on the stiffness of the vertical elements relative to that of the horizontal 
elements and, second, on the relative stiffness of the various vertical elements if they have varying 
deflection characteristics. The first issue is discussed in detail in the Provisions, which define when 
a diaphragm can be considered flexible or rigid and set limits on diaphragms that act in rotation or 
that cantilever. The second is largely an issue of engineering mechanics, but is discussed here 
because significant variations in engineering practice currently exist. 
In situations where a series of vertical elements of the seismic-force-resisting system are aligned in a 
row, seismic forces will distribute to the different elements according to their relative stiffness. 
Typical current design practice is to distribute seismic forces to a line of wood structural panel 
sheathed walls in proportion to the lengths of the wall segments such that each segment carries the 
same unit load. Wood structural panel sheathed wall segments without openings can generally be 
calculated to have a stiffness in proportion to the wall length when: the tie-down slip is ignored, the 
2 003 Commentary, Chapter 12 
242 
wood structural panel sheathing is selected from standard selection tables, and the aspect ratio limits 
of the Provisions are satisfied. For stiffness to be proportional to the wall length, the average load 
per nail for a given nail size must be approximately equal. Conversely, a wall could be stiffened by 
adding nails and reducing the calculated average load per nail. When including tie-down slip from 
anchors with negligible slip (1/16 in. [2 mm] or less), the assumption of wall stiffness proportional 
to length is still fairly reasonable. For larger tie-down slip values, wall stiffness will move towards 
being proportional to the square of the wall length; more importantly, however, the anchorage will 
start exhibiting displacement compatibility problems. For shear walls with aspect ratios higher than 
2/1, the stiffness is no longer in proportion to the length and equations are not available to 
reasonably calculate the stiffness. For a line of walls with variations in tie-down slip, chord framing, 
unit load per nail, or other aspects of construction, distribution of load to wall segments will need to 
be based on a deflection analysis. The shear wall and diaphragm deflection equations that are 
currently available are not always accurate. As testing results become available, the deflection 
calculation formulas will need to be updated and design assumptions for distribution of forces 
reviewed. 
Torsional diaphragm force distribution. A diaphragm is flexible when the maximum lateral 
deformation of the diaphragm is more than two times the average story drift. Conversely, a 
diaphragm will be considered rigid when the diaphragm deflection is equal to or less than two times 
the story drift. This is based on a model building code definition that applies to all materials. 
For flexible diaphragms, seismic forces should be distributed to the vertical force-resisting elements 
according to tributary area or simple beam analysis. Although rotation of the diaphragm may occur 
because lines of vertical elements have different stiffness, the diaphragm is not considered stiff 
enough to redistribute seismic forces through rotation. The diaphragm can be visualized as a singlespan 
beam supported on rigid supports. 
For diaphragms defined as rigid, rotational or torsional behavior is expected and results in 
redistribution of shear to the vertical force-resisting elements. Requirements for horizontal shear 
distribution are in Sec. 5.2.4. Torsional response of a structure due to irregular stiffness at any level 
within the structure can be a potential cause of failure. As a result, dimensional and diaphragm ratio 
limitations are provided for different categories of rotation. Also, additional requirements apply 
when the structure is deemed to have a torsional irregularity in accordance with Table 4.3-2, Item 1a 
or 1b. 
In order to understand limits placed on diaphragms acting in rotation, it is helpful to consider two 
different categories of diaphragms. Category I includes rigid diaphragms that rely on force transfer 
through rotation to maintain stability. An example would be an open front structure with shear walls 
on the other three sides. For this more structurally critical category, applicable limitations are: 
Diaphragm may not be used to resist forces contributed by masonry or concrete in structures over 
one story. 
The length of the diaphragm normal to the opening may not exceed 25 ft ( to perpendicular shear 
walls), and diaphragm L/b ratios are limited as noted. 
Additional limitations apply when rotation is significant enough to be considered a torsional 
irregularity. 
Category II includes rigid diaphragms that have two or more supporting shear walls in each of two 
perpendicular directions but, because the center of mass and center of rigidity do not coincide, 
redistribute forces to shear walls through rotation of the diaphragm. These can be further divided 
into Category IIA where the center of rigidity and mass are separated by a small portion of the 
structures least dimension and the magnitude of the rotation is on the order of the accidental 
rotation discussed in Sec. 5.2.4.2. For this level of rotation, an exception may result in no particular 
limitations being placed on diaphragm rotation for Category IIA. Category IIB, rigid diaphragms 
Wood Structure Design Requirements 
243 
with eccentricities larger than those discussed in Sec. 5.2.4.2, are subject to the following 
limitations: 
Diaphragm may not be used to resist forces contributed by masonry or concrete in structures over 
one story. 
Additional limitations apply when rotation is significant enough to be considered a torsional 
irregularity. 
Because flexible diaphragms have very little capacity for distributing torsional forces, further 
limitation of aspect ratios is used to limit diaphragm deformation such that rigid behavior will occur. 
The resulting deformation demand on the structure also is limited. Where diaphragm ratios are 
further limited, exceptions permit higher ratios where calculations demonstrate that higher 
diaphragm deflections can be tolerated. In this case, it is important to determine the effect of 
diaphragm rigidity on both the horizontal distribution and the ability of other structural elements to 
withstand resulting deformations. 
Proposals to prohibit wood diaphragms acting in rotation were advanced following the 1994 
Northridge earthquake. To date, however, the understanding is that the notable collapses in the 
Northridge earthquake occurred in part because of lack of deformation compatibility between the 
various vertical resisting elements rather than because of the inability of the diaphragm to act in 
rotation. 
Diaphragm cantilever. Limitations concerning diaphragms that cantilever horizontally past the 
outermost shear wall (or other vertical element) are related to but distinct from those imposed 
because of diaphragm rotation. Such diaphragms can be flexible or rigid and for rigid diaphragms 
can be Category I, IIA or IIB. Both the limitations based on diaphragm rotation (if applicable) and 
the following limit on diaphragm cantilever must be considered: 
Diaphragm cantilever may not exceed the lesser of 25 ft or two thirds of the diaphragm width. 
Relative stiffness of vertical elements. In situations where a series of vertical elements of the 
seismic-force-resisting system are aligned in a row, the forces will distribute to the different 
elements according to their relative stiffnesses. This behavior needs to be taken into account 
whether it involves a series of wood structural panel shear walls of different lengths, a mixture of 
wood structural panel shear walls with diagonal lumber or non-wood sheathed shear walls, or a 
mixture of wood shear walls with walls of some other material such as concrete or masonry. 
Diaphragm aspect ratio. The L/b for a diaphragm is intended to be the typical definition for aspect 
ratio. The diaphragm span, L, is measured perpendicular to the direction of applied force, either for 
the full dimension of the diaphragm or between supports as appropriate. The width, b, is parallel to 
the applied force (see Figure C12.3-1). 
Figure C12.3-1 Diaphragm dimension definitions. 
Single and double diagonally sheathed lumber diaphragms. Diagonally sheathed lumber 
diaphragms are addressed by the Provisions because they are still used for new construction in some 
regions. Shear resistance is based on a soft conversion from the model code allowable stress loads 
2 003 Commentary, Chapter 12 
244 
and capacities to Provisions strength loads for regions with high spectral accelerations. This will 
allow users in the western states, where this construction is currently being used, to continue with 
little or no change in requirements; at the same time, reasonable values are provided for regions with 
lower spectral 
Shear wall aspect ratio. The h/b for a shear wall is intended to be the typical definitions for aspect 
ratio. The h of the shear wall is the clear story height (see Figure C12.3-2). The alternate definition 
of aspect ratio is only to be used where specific design and detailing is provided for force transfer 
around the openings. It is required that the individual wall piers meet the aspect ratio requirement 
(see Figure C12.3-3) and that the overall perforated wall also meet the aspect ratio requirement. Use 
of the alternate definition involves the design and detailing of chord and collector elements around 
the opening, and often results in the addition of blocking, strapping, and special nailing. As noted, 
the design for force transfer around the opening must use a rational analysis and be in accordance 
with AF&PA/ASCE 16, which discusses design principles for shear walls, diaphragms, and 
boundary elements. 
In general, unit shear values for wood structural panel sheathing have been based on tests of shear 
wall panels with aspect ratios of 2/1 to 1/1. Narrower wall segments (that is, with aspect ratios 
greater than 2/1) have been a recent concern based on damage observations following the Northridge 
earthquake and based on results of recent research (Applied Technology Council, 1995; White and 
Dolan, 1996). In response, various limitations on aspect ratios have been proposed. In the 
Provisions, an aspect ratio adjustment, 2b/h, is provided to account for the reduced stiffness of 
narrow shear wall segments. This adjustment is based on a review of numerous tests of narrow 
aspect ratio walls by Technical Subcommittee 7. The maximum 3.5/1 aspect ratio is recommended 
based on constructability issues (placement of tie-downs) as well as reduced stiffness of narrower 
shear wall segments. 
Figure C12.3-2 Typical shear wall height-towidth 
ratio. 
Wood Structure Design Requirements 
245 
Figure C12.3-3 Alternate shear wall height-to-width 
ratio with design for force transfer around openings. 
Single and double diagonally sheathed lumber shear walls. Diagonally sheathed lumber shear 
walls are addressed by the Provisions because they are still used for new construction in some 
regions. Resistance values are based on a soft conversion from the model code allowable stress loads 
and capacities to Provisions strength loads for regions with high spectral accelerations. This will 
allow users in the western states, where this construction is currently being used, to continue with 
little or no change in requirements; at the same time, reasonable values are provided for regions with 
lower spectral accelerations. 
Perforated shear walls (PSW). In a traditional engineering approach for design of shear walls with 
openings, design force transfer around the openings involves developing a system of piers and 
coupling beams within the shear wall. Load paths for the shear and flexure developed in the piers 
and coupling beams generally require blocking and strapping extending from each corner of the 
opening to some distance beyond. This approach often results in shear wall detailing that is not 
practical to construct. 
The perforated shear wall approach utilizes empirically based reductions of wood structural panel 
shear wall capacities to account for the presence of openings that have not been specifically 
designed and detailed for moment resistance. This method accounts for the capacity that is inherent 
in standard construction, rather than relying on special construction requirements. It is not expected 
that sheathed wall areas above and below openings behave as coupling beams acting end to end, but 
rather that they provide local restraint at their ends. As a consequence significantly reduced 
capacities are attributed to interior perforated shear wall segments with limited overturning restraint. 
Example 1 and Example 2 provide guidance on the application of the perforated shear wall 
approach. 
2 003 Commentary, Chapter 12 
246 
Perforated Shear Wall Limitations. Perforated shear wall design provisions are applicable to 
wood structural panel shear walls having characteristics identified in this section. 
1. The requirement that perforated shear wall segments be provided at each end of the perforated 
shear wall ensures that a minimum length of full height sheathing, conforming to applicable 
aspect ratio limits, is included at each end of a perforated shear wall. 
2. A factored shear resistance not to exceed 0.64 klf, based on tabulated LRFD values, is provided 
to identify a point beyond which other means of shear wall design are likely to be more 
practical. Connection requirements associated with unadjusted shear resistance greater than 
0.64 klf will likely not be practical as other methods of shear wall design will be more efficient. 
3. Each perforated shear wall segment must satisfy the requirements for shear wall aspect ratios. 
The 2b/h adjustment for calculation of unadjusted factored shear resistance only applies when 
shear wall segments with h/b greater than 2:1 but not exceeding 3.5:1 are used in calculating 
perforated shear wall resistance. When shear wall segments with h/b greater than 2:1 are present 
in a perforated shear wall, but not utilized in calculation of perforated shear wall resistance, 
calculation of unadjusted factored shear resistance should not include the 2b/h adjustment. In 
many cases, due to the conservatism of the 2b/h adjustment, it is advantageous to simply ignore 
the presence of shear wall segments with h/b greater than 2:1 when calculating perforated shear 
wall resistance. 
4. No out-of-plane offsets are permitted in a perforated shear wall. While the limit on out-of-plane 
offsets is not unique to perforated shear walls, it is intended to clearly indicate that a perforated 
shear wall shall not have out-of-plane (horizontal) offsets. 
5. Collectors for shear transfer to each perforated shear wall segment provide for continuity 
between perforated shear wall segments. This is typically achieved through continuity of the 
wall double top plates or by attachment of perforated shear wall segments to a common load 
distributing element such as a floor or roof diaphragm. 
6. Uniform top-of-wall and bottom-of-wall elevations are required for use of the empirical shear 
adjustment factors. 
7. Limiting perforated shear wall height to 20 ft addresses practical considerations for use of the 
method as wall heights greater than 20 ft are uncommon. 
a. The width, L, of a perforated shear wall and widths L1, L2 and L3 of perforated shear 
wall segments are shown in Figure C12.3-4. In accordance with the limitations and 
anchorage requirements, perforated shear wall segments and overturning restraint 
must be provided at each end of the perforated shear wall. 
Perforated shear wall 
L1 L2 L3 
L 
h 
Overturning Restraint (each end) 
Figure C12.3-4 Perforated shear wall. 
Wood Structure Design Requirements 
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Perforated shear wall resistance. Opening adjustment factors are used to reduce shear wall 
resistance, based on the percent full-height sheathing and the maximum opening height ratio. 
Opening adjustment factors are based on the following empirical equation for shear capacity ratio, 
F, which relates the ratio of the shear capacity for a wall with openings to the shear capacity of a 
fully sheathed wall (Sugiyama, 1981): 
4 
3 2 
F 
r 
= 
- 
(C12.3-1a) 
1 
1 o 
i 
r 
h L 
A = 
+ S 
(C12.3-1b) 
where: 
r = sheathing area ratio, 
Ao = total area of openings, 
h = wall height, 
SLi = sum of the width of full-height sheathing. 
Agreement between Eq. C12.3-1a and tabulated opening adjustment factors is achieved by 
recognizing that the tabulated opening adjustment factors are: (1) derived based on an assumption 
that the height of all openings in a wall are equal to the maximum opening height; and, (2) applied to 
the sum of the widths of the shear wall segments meeting applicable height-to-width ratios. The 
assumption that the height of all openings in a wall are equal to the maximum opening height 
conservatively simplifies tabular presentation of shear capacity adjustment factors for walls with 
more than one opening height. 
Early verification of Eq. C12.3-1a was based on testing of one-third and full-scale shear wall 
assemblies (Yasumura, 1984; Sugiyama, 1994). More recently, substantial U.S. verification testing 
of the influence of openings on shear strength and stiffness has taken place (APA, 1996; Dolan and 
Johnson, 1996; Dolan and Heine, 1997; NAHB-RC, 1998) indicating shear wall performance is 
consistent with predictions of Eq. C12.3-1a. Results of cyclic testing indicate that the loss in 
strength due to cyclic loading is reduced for shear walls with openings, indicating good performance 
relative to that of shear walls without openings. Figure C12.3-5 provides a graphical summary of 
some recent U.S. verification testing. Data from monotonic tests of 12-ft shear walls (APA, 1996), 
monotonic and cyclic tests of long shear walls with unsymmetrically placed openings (Dolan and 
Johnson, 1996), and monotonic and cyclic tests of 16-ft and 20-ft shear walls with narrow wall 
segments (NAHB-RC, 1998). 
2 003 Commentary, Chapter 12 
248 
Sheathing Area Ratio, r 
0.0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0 
Shear 
Capacity 
Ratio, F 
0.0 
0.1 
0.2 
0.3 
0.4 
0.5 
0.6 
0.7 
0.8 
0.9 
1.0 
F=r/(3-2r) 
APA - 1996 
Dolan - 1996 
NAHB-RC -1998 
Figure C12.3-5 Shear capacity ratios (actual and predicted). 
Eq. C12.3-1a for shear load ratio, F, has been shown to be a good approximation of the stiffness 
ratio of a wall with openings to that of a fully sheathed wall. Accordingly, the deflection of a 
perforated shear wall can be calculated as the deflection of an equivalent length fully sheathed wall, 
divided by the shear load ratio, F. 
The percent full-height sheathing and the maximum opening height ratio are used to determine an 
opening adjustment factor. Maximum opening height is the maximum vertical dimension of an 
opening within the perforated shear wall. A maximum opening height equal to the wall height is 
used where structural sheathing is not present above or below window openings or above door 
openings. The percent full-height sheathing is calculated as the sum of the widths of perforated 
shear wall segments divided by the total width of the shear wall. Sections sheathed full-height 
which do not meet aspect ratio limits for wood structural panel shear walls are not considered in 
calculation of percent full-height sheathing. 
PSW Anchorage and load path. Anchorage for uplift at perforated shear wall ends, shear, uplift 
between perforated shear wall ends, and compression chord forces are prescribed to address the nonuniform 
distribution of shear within a perforated shear wall. 
Prescribed forces for shear and uplift connections ensure that the capacity of the wall is governed by 
the sheathing to framing attachment (shear wall nailing) and not bottom plate attachment for shear 
and/or uplift. Shear and uplift forces approach the unadjusted factored shear resistance of the 
perforated shear wall segment as the shear load approaches the shear resistance of the perforated 
shear wall. A continuous load path to the foundation based on this requirement and consideration of 
other forces (for example, from the story above) shall be maintained. The magnitude of shear and 
uplift varies as a function of overturning restraint provided and aspect ratio of the shear wall 
segment. 

Wood Structure Design Requirements 
249 
Uplift anchorage at perforated shear wall ends. Anchorage for uplift forces due to overturning 
are required at each end of the perforated shear wall. A continuous load path to the foundation 
based on this requirement and consideration of other forces (for example, from the story above) shall 
be maintained. In addition, compression chords of perforated shear wall segments are required to 
transmit compression forces equal to the required tension chord uplift force. 
PSW Anchorage for in-plane shear. It is required that fastening be provided along the length of 
the sill plate of wall sections sheathed full-height to resist distributed shear, v, and uplift, t, forces. 
The resistance required for the shear connection is the average shear over the perforated shear wall 
segments, divided by the adjustment factor. This resistance will approach the unadjusted factored 
shear resistance of the wall as the shear wall demand approaches the maximum resistance. This 
shear fastening resistance conservatively accounts for the non-uniform distribution of shear within a 
perforated shear wall, since it represents the shear that can only be achieved when full overturning 
restraint is provided. 
Provisions require that distributed fastening for shear, v, and uplift, t, be provided over the length of 
full-height sheathed wall sections. With no other specific requirements, the fastening between the 
full height segments will be controlled by minimum construction fastening requirements. For 
bottom plates on wood platforms this would only require one 16-penny nail at 16 in. on center. In 
some cases, it may be preferable to extend a single bottom plate fastening schedule across the entire 
length of the perforated shear wall rather than to require multiple fastening schedules. 
Uplift anchorage between perforated shear wall ends. The resistance required for distributed 
uplift anchorage, t, is the same as the required shear resistance, v. The adequacy of the distributed 
uplift anchorage can be demonstrated using principles of mechanics and recent testing that 
determined the capacity of shear wall segments without uplift anchorage. A 4-ft wide shear wall 
segment with distributed anchorage of the base plate in lieu of an uplift anchor device provided 
about 25 percent of the resistance of a segment with uplift anchorage; an 8-ft wide shear wall 
segment resisted about 45 percent. When these are combined with the resistance adjustment factors, 
overturning resistance based on the unadjusted factored shear resistance is adequate for perforated 
shear wall segments with full height openings on each side. Conceptually the required distributed 
uplift resistance is intended to provide the same resistance that anchor bolts spaced at 2 ft on center 
provided for tested assemblies. While in the tested assemblies the bottom plates were fastened 
down, for design it is equally acceptable to fasten down the studs with a strap or similar device, 
since the studs will in turn restrain the bottom plate. 
PSW Load path. A continuous load path to the foundation is required for the uplift resistance, T; 
the compression resistance, C; the unit shear resistance, v; and the unit uplift resistance, t. 
Consideration of accumulated forces (for example, from the stories above) is required. Where shear 
walls occur at the same location at each floor (stack), accumulation of forces is reasonably 
straightforward. Where shear walls do not stack, attention will need to be paid to maintaining a load 
path for tie-downs at each end of the perforated shear wall, for compression resistance at each end of 
each perforated shear wall segment, and for distributed forces v and t at each perforated shear wall 
segment. Where ends of shear perforated shear wall segments occur over beams or headers, the 
beam or header will need to be checked for the vertical tension and compression forces in addition to 
gravity forces. Where adequate collectors are provided at lower floor shear walls, the total shear 
wall load need only consider the average shear in the perforated shear wall segments above, and not 
the average shear divided by the adjustment factor. 

2 003 Commentary, Chapter 12 
250 

Wood Structure Design Requirements 
251 

2 003 Commentary, Chapter 12 
252 
Wood Structure Design Requirements 
253 
Figure C12.3-6 Elevation for perforated shear wall Example 2.
2 003 Commentary, Chapter 12 
254 
Figure C12.3-7 Details for perforated shear wall Example 2.
Wood Structure Design Requirements 
255 
12.3.1 Framing. All framing that is designed as part of an engineered wood structure must be 
designed with connectors that are able to transfer the required forces between various components. 
These connectors can be either proprietary hardware or some of the more conventional connections 
used in wood construction. However, these connectors should be designed according to accepted 
engineering practice to ensure that they will have the capacity to resist the forces. The requirement of 
columns and posts being framed to full end bearing requires that the force transfer from the column to 
the base be accomplished through end grain bearing of the wood, not through placing the bolts or other 
connectors in shear. This requirement is included to ensure adequate capacity for transfer of the 
vertical forces due to both gravity and overturning moment. Alternatively, the connection can be 
designed to transfer the full loading through placing the bolts or other connectors in shear neglecting 
all possible bearing. 
The anchorage connections used in engineered wood construction must be capable of resisting the 
forces that will occur between adjacent members (beams and columns) and elements (diaphragms and 
shear walls). These connections can utilize proprietary hardware or be designed in accordance with 
principles of mechanics. Inadequate connections are often the cause of structural failures in wood 
structures, and the registered design professional is cautioned to use conservative values for allowable 
capacities since most published values are based on monotonic, not cyclic, load applications (U.S. 
Department of Agriculture, National Oceanic and Atmospheric Administration, 1971). Testing has 
shown that some one-sided bolted connections subject to cyclic loading, such as tie-down devices, do 
not perform well. This was substantiated by the poor performance of various wood frame elements in 
structures in the January 1994 Northridge earthquake. 
Concrete or masonry wall anchorages using toe nails or nails subject to withdrawal are prohibited by 
the Provisions. It has been shown that these types of connections are inadequate and do not perform 
well (U.S. Department of Agriculture, National Oceanic and Atmospheric Administration, 1971). 
Ledgers subjected to cross-grain bending or tension perpendicular to grain also have performed poorly 
in past earthquakes, and their use is now prohibited by the Provisions. 
12.4 CONVENTIONAL LIGHT-FRAME CONSTRUCTION 
The Provisions intend that a structure using conventional construction methods and complying with 
the requirements of this section be deemed capable of resisting the seismic forces imposed by the 
Provisions. 
Repetitive framing members such as joists, rafters, and studs together with sheathing and finishes 
comprise conventional light-frame construction. The subject of conventional construction is addressed 
in each of the model codes. It is acknowledged and accepted that, for the most part, the conventional 
construction provisions in the model codes concerning framing members and sheathing that carry 
gravity loads are adequate. This is due to the fact that the tables in the model codes giving allowable 
spans have been developed using basic principles of mechanics. For seismic lateral force resistance, 
however, experience has shown that additional requirements are needed. 
To provide lateral force resistance in vertical elements of structures, wall bracing requirements have 
been incorporated in conventional construction provisions of the model codes. With a few exceptions, 
these generally have been adequate for single family residences for which conventional construction 
requirements were originally developed. While the model building codes have been quite specific as 
to the type of bracing materials to be used and the amount of bracing required in any wall, no limits on 
the number or maximum separation between braced walls have been established. This section of the 
Provisions introduces the concept of mandating the maximum spacing of braced wall lines. By 
mandating the maximum spacing of braced wall lines and thereby limiting the lateral forces acting on 
these vertical elements, these revisions provide for a seismic-force-resisting system that will be less 
prone to overstressing and the requirements can be applied and enforced more uniformly than previous 
model building code requirements. While specific elements of light-frame construction may be 
calculated to be overstressed, there is typically a great deal of redundancy and uncounted resistance in 
such structures and they have generally performed well in past earthquakes. The experience in the 
2 003 Commentary, Chapter 12 
256 
Northridge earthquake was, however, less reassuring, especially for those residences relying on 
gypsum board or stucco for lateral force resistance. The light weight of conventional construction, 
together with the large energy dissipation capacity of the multiple fasteners used and inherent 
redundancy of the system are major factors in the observed good performance where wood or woodbased 
panels were used. 
12.4.1 Limitations 
12.4.1.1 General. The scope of this section specifically excludes prescriptive design of structures 
with concrete or masonry walls above the basement story, with the exception of veneer, in order to 
maintain the light weight of construction that the bracing requirements are based on. Wood braced 
wall panels and diaphragms as prescribed in this section are not intended to support lateral forces due 
to masonry or concrete construction. Prescriptive (empirical) design of masonry walls is allowed for 
in Chapter 11; however, design of structures combining masonry wall construction and wood roof and 
floor diaphragm construction must have an engineered design. In regions of high seismic activity, past 
earthquakes have demonstrated significant problems with structures combining masonry and wood 
construction. While engineered design requirements do address these problems, the prescriptive 
requirements in the model codes do not adequately address these problems. Masonry and concrete 
basement walls are permitted to be constructed in accordance with the requirements of the IRC. 
12.4.1.2 Irregular structures. This section was added to the 1997 Provisions to clarify the definition 
of irregular (unusually shaped) structures that would require the structure to be designed for the forces 
prescribed in Chapter 5 in accordance with the requirements of Sec. 12.3 and 12.4. The descriptions 
and diagrams provide the registered design professional with several typical irregularities that produce 
torsional response, or result in forces considered high enough to require an engineered design and 
apply only to structures assigned to Seismic Design Category C or D. 
Structures with geometric discontinuities in the lateral-force-resisting system have been observed to 
sustain more earthquake and wind damage than structures without discontinuities. They have also 
been observed to concentrate damage at the discontinuity location. For Seismic Design Categories C 
and D, this section translates applicable irregularities from Tables 4.3-2 and 4.3-3 into limitations on 
conventional light-frame construction. If the described irregularities apply to a given structure, it is 
required that either the entire structure or the non-conventional portions be engineered in accordance 
with the engineered design portions of the Provisions. The irregularities are based on similar model 
code requirements. While conceptually these are equally applicable to all seismic design categories, 
they are more readily accepted in areas of high seismic risk, where damage due to irregularities has 
been observed repeatedly. 
Application of engineered design to non-conventional portions rather than to the entire structure is a 
common practice in some regions. The registered design professional is left to judge the extent of the 
portion to be designed. This often involves design of the nonconforming element, force transfer into 
the element, and a load path from the element to the foundation. A nonconforming portion will 
sometimes have enough of an impact on the behavior of a structure to warrant that the entire seismicforce-
resisting system receives an engineered design. 
12.4.1.2.1 Out-of-plane offset. This limitation is based on Item 4 of Table 4.3-2 and applies when 
braced wall panels are offset out-of-plane from floor to floor. In-plane offsets are discussed in another 
item. Ideally braced wall panels would always stack above of each other from floor to floor with the 
length stepping down at upper floors as less length of bracing is required. 
Because cantilevers and set backs are very often incorporated into residential construction, the 
exception offers rules by which limited cantilevers and setbacks can be considered conventional. 
Floor joists are limited to 2 by 10 (actual: 12 by 93 in.; 38 by 235 mm) or larger and doubled at braced 
wall panel ends in order to accommodate the vertical overturning reactions at the end of braced wall 
panels. In addition the ends of cantilevers are attached to a common rim joist to allow for 
redistribution of load. For rim joists that cannot run the entire length of the cantilever, the metal tie is 
Wood Structure Design Requirements 
257 
intended to transfer vertical shear as well as to provide a nominal tension tie. Limitations are placed 
on gravity loads to be carried by cantilever or setback floor joists so that the joist strength will not be 
exceeded. The roof loads discussed are based on the use of solid sawn members where allowable 
spans limit the possible loads. Where engineered framing members such as trusses are used, gravity 
load capacity of the cantilevered or setback floor joists should be carefully evaluated. 
12.4.1.2.2 Unsupported diaphragm. This limitation is based in Item 1 of Table 4.3-2, and applies to 
open-front structures or portions of structures. The conventional construction bracing concept is based 
on using braced wall lines to divide a structure up into a series of boxes of limited dimension, with the 
seismic force to each box being limited by the size. The intent is that each box be supported by braced 
wall lines on all four sides, limiting the amount of torsion that can occur. The exception, which 
permits portions of roofs or floors to extend past the braced wall line, is intended to permit 
construction such as porch roofs and bay windows. Walls for which lateral resistance is neglected are 
allowed in areas where braced walls are not provided. 
12.4.1.2.3 Opening in wall below. This limitation is based on Item 4 of Table 4.3-3 and applies 
when braced wall panels are offset in-plane. Ends of braced wall panels supported on window or door 
headers can be calculated to transfer large vertical reactions to headers that may not be of adequate 
size to resist these reactions. The exception permits a 1 ft extension of the braced wall panel over a 4 
by 12 (actual: 32 by 113 in.; 89 by 286 mm) header on the basis that the vertical reaction is within a 45 
degree line of the header support and therefore will not result in critical shear or flexure. All other 
header conditions require an engineered design. Walls for which lateral resistance is neglected are 
allowed in areas where braced walls are not provided. 
12.4.1.2.4 Vertical offset in diaphragm. This limitation results from observation of damage that is 
somewhat unique to split-level wood frame construction. If floors on either side of an offset move in 
opposite directions due to earthquake or wind loading, the short bearing wall in the middle becomes 
unstable and vertical support for the upper joists can be lost, resulting in a collapse. If the vertical 
offset is limited to a dimension equal to or less than the joist depth, then a simple strap tie directly 
connecting joists on different levels can be provided, eliminating the irregularity. The IRC, Sec. 
502.6.1, provides requirements for tying of floor joists. 
12.4.1.2.5 Non-perpendicular walls. This limitation is based on Item 5 of Table 4.3-2 and applies to 
nonperpendicular braced wall lines. When braced wall lines are not perpendicular to each other, 
further evaluation is needed to determine force distributions and required bracing. 
12.4.1.2.6 Large diaphragm opening. This limitation is based on Item 3 of Table 4.3-2 and attempts 
to place a practical limit on openings in floors and roofs. Because stair openings are essential 
toresidential construction and have long been used without any report of life-safety hazards resulting, 
these are felt to be acceptable conventional construction. See Sec. 12.4.3.7 for detailing requirements 
for permitted openings. 
12.4.1.2.7 Stepped foundation. This limits a condition that can cause a torsional irregularity per 
Item 1 of Table 4.3-2. Where heights of braced wall panels vary significantly, distribution of lateral 
forces will also vary. If a structure on a hill is supported on 2-ft-high, braced cripple wall panels on 
one side and 8-ft-high panels on the other, torsion and redistribution of forces will occur. An 
engineered design for this situation is required in order to evaluate force distribution and provide 
adequate wall bracing and anchor bolting. This limitation applies specifically to walls from the 
foundation to the floor. While gable-end walls have similar variations in wall heights, this has not 
been observed to be a significant concern in conventional construction. See Sec. 12.4.3.6 for detailing 
requirements for permitted foundation stepping. 
12.4.2 Braced walls 
12.4.2.1 Spacing between braced wall lines. Table 12.4-1 prescribes the spacing of braced wall lines 
and number of stories permitted for conventional construction structures. Figures C12.4-1 and C12.4-
2 003 Commentary, Chapter 12 
258 
2 illustrate the basic components of the lateral bracing system. Information in Tables 12.4-1 and 12.4- 
2 was first included in the 1991 Provisions. 
12.4.2.2 Braced wall line sheathing. Table 12.4-2 prescribes the minimum length of bracing along 
each 25 ft (7.6 m) length of braced wall line. Total height of structures has been reduced to limit 
overturning of the braced walls so that significant uplift is not generally encountered. The height limit 
will accommodate 8 to 10 ft (2.4 to 3 m) story heights. 
12.4.3 Detailing requirements. The intent of this section is to rely on the traditional light-frame 
conventional construction materials and fastenings as prescribed in the references for this chapter. 
Braced wall panels are not required to be aligned vertically or horizontally (within the limits 
prescribed in Sec. 12.4.1) but stacking is desirable where possible. With the freedom provided for 
non-alignment it becomes important that a load path be provided to transfer lateral forces from upper 
levels through intermediate vertical and horizontal resisting elements to the foundation. Connections 
between horizontal and vertical resisting elements are prescribed. In structures two or three stories in 
height, it is desirable to have interior braced wall panels supported on a continuous foundation. See 
Figures C12.4-3 through C12.4-13 for examples of connections. 
The 1997 Provisions incorporated some of the wall anchorage, top plate, and braced wall panel 
connection requirements from the model building codes. These are included for completeness of the 
document and to clarify the requirement for the registered design professional. Additional 
requirements for foundations supporting braced wall panels has also been added to provide guidance 
and clarity for the registered design professional. 

Wood Structure Design Requirements 
259 
Figure C12.4-1 Acceptable one-story bracing example. 

2 003 Commentary, Chapter 12 
260 
Figure C12.4-2 Acceptable two-story bracing example. 

Wood Structure Design Requirements 
261 
Figure C12.4-3 Wall anchor detail. 
2 003 Commentary, Chapter 12 
262 
Figure C12.4-4 Double top splice. 
Figure C12.4-5 Single top splice. 

Wood Structure Design Requirements 
263 
Figure C12.4-6 Full bearing bottom plate. 
Figure C12.4-7 Exterior braced wall.
2 003 Commentary, Chapter 12 
264 
Figure C12.4-8 Interior braced wall at perpendicular joist. 
Figure C12.4-9 Interior braced wall at parallel joist. 

Wood Structure Design Requirements 
265 
Figure C12.4-10 Offset at interior braced wall. 

2 003 Commentary, Chapter 12 
266 
Figure C12.4-11 Diaphragm connection to braced wall below 
Figure C12.4-12 Post base detail. 
Wood Structure Design Requirements 
267 
Figure C12.4-13 Wood beam connection to post. 
12.4.3.4 Braced wall panel connections. The exception provided in this section of the Provisions is 
included due to the difficulty in providing a mechanism to transfer the diaphragm loads from a truss 
roof system to the braced wall panels of the top story. This problem has been considered by the 
Clackamas County, Oregon Building Codes Division, and an alternate to the CABO Building Code 
Sec. 402.10 was written in 1993, and revised September 5, 1995. The details shown in Figure C12.4- 
14 through C12.4-17 are provided as suggested methods for providing positive transfer of the lateral 
forces from the diaphragm through the web sections of the trusses to the top of the braced wall panels 
below. 
Figure C12.4-14 Suggested methods for transferring roof diaphragm load s to braced wall 
panels. 

2 003 Commentary, Chapter 12 
268 
Figure C12.4-15 Alternate gable end brace. 
Figure C12.4-16 Wall parallel to truss bracing detail. 

Wood Structure Design Requirements 
269 
Figure C12.4-17 Wall parallel to truss alternate bracing detail. 
REFERENCES 
American Institute of Timber Construction. 1994. Timber Construction Manual. New York, New 
York: John Wiley and Sons, Inc. 
American Forest & Paper Association. 1991. ANSI/NFoPA NDS-1991. National Design 
Specification for Wood Construction. Washington, D.C.: AF&PA. 
American Forest & Paper Association. 1996. Manual of Wood Construction: Load and Resistance 
Factor Design (LRFD). Washington, D.C.: AF&PA. 
American Plywood Association. 1991. Design/Construction GuideDiaphragms. Tacoma, 
Washington: APA. 
American Plywood Association. 1994. Northridge California Earthquake, T94-5. Tacoma, 
Washington.: APA. 
American Plywood Association. 1996. Proposed Cyclic Testing Standard for Shear Walls. Tacoma, 
Washington: APA. 
APA B The Engineered Wood Association. 1996. Wood structural panel shear walls with gypsum 
wallboard and window/door openings, Report 157. Tacoma, Washington: APA. 
ASCE Committee on Wood Load and Resistance Factor Design for Engineered Wood Construction. 
1988. A Pre Standard Report American Society of Civil Engineers. New York, New York: ASCE. 
ASCE. 1995. Standard For Load and Resistance Factor For Engineered Wood Construction. 
AF&PA/ASCE 16-95. American Society of Civil Engineers, New York, New York. 
Applied Technology Council. 1981. Guidelines for the Design of Horizontal Wood Diaphragms, 
ATC-7. Redwood City, California: ATC. 
2 003 Commentary, Chapter 12 
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Applied Technology Council. 1995. Cyclic Testing of Narrow Plywood Shear Walls, ATC-R-1. 
Redwood City, California: ATC. 
Breyer, Donald E. 1993. Design of Wood Structures, Third Edition. New York, New York: 
McGraw-Hill Book Company. 
Canadian Wood Council. 1990. Wood Design Manual. Ottawa: Canadian Wood Council. 
Canadian Wood Council. 1991. Wood Reference Handbook. Ottawa: Canadian Wood Council. 
Council of American Building Officials (CABO). 1995. One- and Two-Family Dwelling Code. 
Country Club Hills, Illinois: BOCA 
Department of the Army, Navy and Air Force. 1992. Seismic Design for Buildings, TM5-809-10 
(Tri-Services Manual). Washington, D.C.: U.S. Government Printing Office. 
Dolan, J.D., 1994. Proposed Test Method for Dynamic Properties of Connections Assembled with 
Mechanical Fasteners. ASTM Journal of Testing and Evaluation 22(6):542-547. 
Dolan, J. D., and M. W. White. 1992. Design Considerations for Using Adhesives in Shear Walls. 
ASCE Journal of Structural Engineering 118(12):3473-3480. 
Dolan, J. D. and Johnson, A. C. 1996. Monotonic Tests of Long Shear Walls with Openings 
Virginia Polytechnic Institute and State University, Timber Engineering Report No. TE-1996-001. 
Blacksburg, Virginia: VPISU. 
Dolan, J. D. and Johnson, A. C. 1996. Cyclic Tests of Long Shear Walls with Openings Virginia 
Polytechnic Institute and State University, Timber Engineering Report No. TE-1996-002. Blacksburg, 
Virginia: VPISU. 
Dolan, J. D., and C. P. Heine. 1997a. Monotonic tests of wood-frame shear walls with various 
openings and base restraint configurations, Timber Engineering Report TE-1997-001. Blacksburg: 
Virginia Polytechnic Institute and State University. 
Dolan, J. D., and C. P. Heine. 1997b. Sequential phased displacement cyclic tests of wood-frame 
shear walls with various openings and base restraint configurations, Timber Engineering Report TE- 
1997-002. Blacksburg: Virginia Polytechnic Institute and State University. 
Earthquake Engineering Research Institute. 1996. Northridge earthquake reconnaissance report, 
Chapter 6, Supplement C to Volume 11, pp. 125 et seq., Earthquake Spectra. 
Faherty, Keith F., and T. G. Williamson. 1989. Wood Engineering and Construction Handbook. 
New York, New York: McGraw-Hill. 
U.S. Department of Housing and Urban Development, 1984. HUD Minimum Property Standards, 
Vol. I, II, and III. Washington, D.C.: United States Government Printing Office. 
Forest Products Laboratory. 1986. Wood: Engineering Design Concepts. University Park: Materials 
Education Council, The Pennsylvania State University. 
Foschi, R. O., and A. Filiatrault. 1990. Performance Evaluation of 3M Scotch Grip Wood Adhesive 
5230 for the Static and Dynamic Design of Timber Shear Walls and Diaphragms, Vancouver: 
Department of Civil Engineering Report, University of British Columbia. 
Goetz, Karl-Heinz, Dieter Hoor, Karl Moehler, and Julius Natterer. 1989. Timber Design and 
Construction Source Book: A Comprehensive Guide to Methods and Practice. New York, New York: 
McGraw-Hill. 
Hoyle and Woeste. 1989. Wood Technology and Design of Structures. Iowa State University Press. 
International Conference of Building Officials (ICBO). 1994. Uniform Building Code. Whittier, 
California: ICBO. 
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Keenan, F. J. 1986. Limit States Design of Wood Structures. Morrison Hershfield Limited. 
NAHB Research Center. 1998. The performance of shear walls with narrow wall segments, reduced 
based restraint, and alternative framing methods. Upper Marlboro, Maryland. 
Rose, J. D., 1996. Research Report 158. Preliminary Testing of Wood Structural Panel Shear Walls 
Under Cyclic (Reversed) Loading. Tacoma, Washington: APA. 
Sherwood and Stroh. 1989. Wood-Frame House Construction in Agricultural Handbook 73. 
Washington, D.C.: U.S. Government Printing Office. 
Somayaji, Shan. 1992. Structural Wood Design. St. Paul, Minnesota: West Publishing Co. 
Stalnaker, Judith J., and E. C. Harris. 1996. Structural Design in Wood, Second Edition. New York, 
New York: McGraw-Hill. 
Sugiyama, H. 1981. The evaluation of shear strength of plywood-sheathed walls with openings, 
Wood Industry, 36-7, July. 
Sugiyama, H., and T. Matsumoto. 1994. Empirical equations for the estimation of racking strength 
of plywood-sheathed shear wall with openings, in Summaries of Technical Papers of Annual 
Meeting, Architectural Institute of Japan, September. 
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Fernando, California, Earthquake of February 9, 1971. Washington, D.C.: NOAA. 
White, M. W. and J. D. Dolan, 1996. Seismic Response of Timber Shear Walls. Part I: Aspect 
Ratios. Paper submitted for publication in ASCE Journal of Structural Engineering. 
Yasumura, M., and H. Sugiyama. 1984. Shear properties of plywood-sheathed wall panels with 
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Additional references are listed in John Peterson. 1983. Bibliography on Lumber and Wood Panel 
Diaphragms, Journal of Structural Engineering, Vol. 109 No. 12. American Society of Civil 
Engineers, New York, New York. 

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 273 
Chapter 13 Commentary 
SEISMICALLY ISOLATED STRUCTURE DESIGN REQUIREMENTS 
13.1 GENERAL 
Seismic isolation, commonly referred to as base isolation, is a design concept based on the premise that 
a structure can be substantially decoupled from potentially damaging earthquake motions. By 
substantially decoupling the structure from the ground motion, the level of response in the structure can 
be reduced significantly from the level that would otherwise occur in a conventional, fixed-base 
building. . 
The potential advantages of seismic isolation and the recent advancements in isolation-system products 
already have led to the design and construction of over 200 seismically isolated buildings and bridges in 
the United States. A significant amount of research, development, and application activity has occurred 
over the past 20 years. The following references provide a summary of some of the work that has been 
performed: Applied Technology Council (ATC, 1986, 1993, and 2002), ASCE Structures Congress 
(ASCE, 1989, 1991, 1993, and 1995), EERI Spectra (EERI, 1990), Skinner, et al. (1993), U.S. 
Conference on Earthquake Engineering (1990 and 1994), and World Conference on Earthquake 
Engineering (1988, 1992, 1996, and 2000). 
In the mid-1980s, the initial applications identified a need to supplement existing codes with design 
requirements developed specifically for seismically isolated buildings. Code development work 
occurred throughout the late 1980s. The status of U.S. seismic isolation design requirements as of May 
2003 is as follows: 
1. In late 1989, the Structural Engineers Association of California (SEAOC) State Seismology 
Committee adopted an Appendix to Chapter 2 of the SEAOC Blue Book entitled, General 
Requirements for the Design and Construction of Seismic-Isolated Structures. These requirements 
were submitted to the International Conference of Building Officials (ICBO) and were adopted by 
ICBO as an appendix of the 1991 Uniform Building Code (UBC). The most current version of these 
regulations may be found in the ASCE-7-02 (ASCE, 2003) and the 2003 International Building 
Code (ICC, 2003).. 
2. In 1991 the Federal Emergency Management Agency (FEMA) initiated a 6-year program to develop 
a set of nationally applicable guidelines for seismic rehabilitation of existing buildings. These 
guidelines (known as the NEHRP Guidelines for the Seismic Rehabilitation of Buildings) were 
published as FEMA 273. In 2000, FEMA 273 was republished, with minor amendments, as FEMA 
356, Prestandard and Commentary for the Seismic Rehabilitation of Buildings. The design and 
analysis methods of the NEHRP Guidelines and the FEMA Prestandard parallel closely methods 
required by the NEHRP Recommended Provisions for new buildings, except that more liberal 
design is permitted for the superstructure of a rehabilitated building. 
A general concern has long existed regarding the applicability of different types of isolation systems. 
Rather than addressing a specific method of base isolation, the Provisions provides general design 
requirements applicable to a wide range of possible seismic isolation systems. 
Although remaining general, the design requirements rely on mandatory testing of isolation-system 
hardware to confirm the engineering parameters used in the design and to verify the overall adequacy of 
the isolation system. Some systems may not be capable of demonstrating acceptability by test and, 
consequently, would not be permitted. In general, acceptable systems will: (1) remain stable for
2003 Commentary, Chapter 13 
274 
required design displacements, (2) provide increasing resistance with increasing displacement, (3) not 
degrade under repeated cyclic load, and (4) have quantifiable engineering parameters (such as forcedeflection 
characteristics and damping). 
Conceptually, there are four basic types of isolation system force-deflection relationships. These 
idealized relationships are shown in Figure C13.1-1 with each idealized curve having the same design 
displacement, DD, for the design earthquake. A linear isolation system is represented by Curve A and 
has the same isolated period for all earthquake load levels. In addition, the force generated in the 
superstructure is directly proportional to the displacement across the isolation system. 
A hardening isolation system is represented by Curve B. This system is soft initially (long effective 
period) and then stiffens (effective period shortens) as the earthquake load level increases. When the 
earthquake load level induces displacements in excess of the design displacement in a hardening system, 
the superstructure is subjected to higher forces and the isolation system to lower displacements than a 
comparable linear system. 
A softening isolation system is represented by Curve C. This system is stiff initially (short effective 
period) and softens (effective period lengthens) as the earthquake load level increases. When the 
earthquake load level induces displacements in excess of the design displacement in a softening system, 
the superstructure is subjected to lower forces and the isolation system to higher displacements than a 
comparable linear system. 
A sliding isolation system is represented by Curve D. This system is governed by the friction force of 
the isolation system. Like the softening system, the effective period lengthens as the earthquake load 
level increases and loads on the superstructure remain constant. 
The total system displacement for extreme displacement of the sliding isolation system, after repeated 
earthquake cycles, is highly dependent on the vibratory characteristics of the ground motion and may 
exceed the design displacement, DD. Consequently, minimum design requirements do not adequately 
define peak seismic displacement for seismic isolation systems governed solely by friction forces. 
13.1.1 Scope. The requirements of Chapter 13 provide isolator design displacements, shear forces for 
structural design, and other specific requirements for seismically isolated structures. All other design 
requirements including loads (other than seismic), load combinations, allowable forces and stresses, and 
horizontal shear distribution are covered by the applicable sections of the Provisions for conventional, 
fixed-base structures. 
Figure C13.1-1 Idealized force-deflection relationships for 
isolation systems (stiffness effects of sacrificial windrestraint 
systems not shown for clarity). 
Seismically Isolated Structure Design Requirements 
275 
13.2 GENERAL DESIGN REQUIREMENTS 
13.2.1 Occupancy importance factor. Ideally, most of the lateral displacement of an isolated 
structure will be accommodated by deformation of the isolation system rather than distortion of the 
structure above. Accordingly, the lateral-load-resisting system of the structure above the isolation 
system should be designed to have sufficient stiffness and strength to avoid large, inelastic 
displacements. For this reason, the Provisions contains criteria that limit the inelastic response of the 
structure above the isolation system. Although damage control for the design-level earthquake is not an 
explicit objective of the Provisions, an isolated structure designed to limit inelastic response of the 
structural system also will reduce the level of damage that would otherwise occur during an earthquake. 
In general, isolated structures designed in conformance with the Provisions should be able: 
1. To resist minor and moderate levels of earthquake ground motion without damage to structural 
elements, nonstructural components, or building contents; and 
2. To resist major levels of earthquake ground motion without failure of the isolation system, without 
significant damage to structural elements, without extensive damage to nonstructural components, 
and without major disruption to facility function. 
The above performance objectives for isolated structures considerably exceed the performance 
anticipated for fixed-base structures during moderate and major earthquakes. Table C13.2-1 provides a 
tabular comparison of the performance expected for isolated and fixed-base structures designed in 
accordance with the Provisions. Loss of function is not included in Table C13.2-1. For certain (fixedbase) 
facilities, loss of function would not be expected to occur until there is significant structural 
damage causing closure or restricted access to the building. In other cases, the facility could have only 
limited or no structural damage but would not be functional as a result of damage to vital nonstructural 
components and contents. Isolation would be expected to mitigate structural and nonstructural damage 
and to protect the facility against loss of function. 
Table C13.2-1 Protection Provided by NEHRP Recommended Provisions for Minor, 
Moderate, and Major Levels of Earthquake Ground Motion 
Earthquake Ground Motion Level 
Risk Category 
Minor 
Moderate 
Major 
Life safetya 
F, I 
F, I 
F, I 
Structural damageb 
F, I 
F, I 
I 
Nonstructural damagec (contents damage) 
F, I 
I 
I 
a Loss of life or serious injury is not expected. 
b Significant structural damage is not expected. 
c Significant nonstructural (contents) damage is not expected. 
F indicates fixed base; I indicates isolated. 
13.2.3.1 Design spectra. Site-specific design spectra must be developed for both the design earthquake 
and the maximum considered earthquake if the structure is located at a site with S1 greater than 0.60 or 
on a Class F site. All requirements for spectra are in Sec. 3.3 and 3.4. 
13.2.4 Procedure selection. The design requirements permit the use of one of three different analysis 
procedures for determining the design-level seismic loads. The first procedure uses a simple, lateralforce 
formula (similar to the lateral-force coefficient now used in conventional building design) to 
prescribe peak lateral displacement and design force as a function of spectral acceleration and isolatedbuilding 
period and damping. The second and third methods, which are required for geometrically 
complex or especially flexible buildings, rely on dynamic analysis procedures (either response spectrum 
or time history) to determine peak response of the isolated building. 
2003 Commentary, Chapter 13 
276 
The three procedures are based on the same level of seismic input and require a similar level of 
performance from the building. There are benefits in performing a more complex analysis in that 
slightly lower design forces and displacements are permitted as the level of analysis becomes more 
sophisticated. The design requirements for the structural system are based on the design earthquake, a 
severe level of earthquake ground motion defined as two-thirds of the maximum considered earthquake. 
The isolation systemincluding all connections, supporting structural elements, and the gapis 
required to be designed (and tested) for 100 percent of maximum considered earthquake demand. 
Structural elements above the isolation system are not required to be designed for the full effects of the 
design earthquake , but may be designed for slightly reduced loads (that is, loads reduced by a factor of 
up to 2.0) if the structural system has sufficient ductility, etc., to respond inelastically without sustaining 
significant damage. A similar fixed-base structure would be designed for loads reduced by a factor of 8 
rather than 2. 
This section delineates the requirements for the use of the equivalent lateral force procedure and 
dynamic methods of analysis. The limitations on the simplified lateral-force design procedure are quite 
severe at this time. Limitations cover the site location with respect to active faults; soil conditions of the 
site, the height, regularity and stiffness characteristics of the building; and selected characteristics of the 
isolation system. Response-history analysis is required to determine the design displacement of the 
isolation system (and the structure above) for the following isolated structures: 
1. Isolated structures with a nonlinear isolation system including, but not limited to, isolation 
systems utilizing friction or sliding surfaces, isolation systems with effective damping values greater 
than about 30 percent of critical, isolation systems not capable of producing a significant restoring 
force, and isolation systems that restrain or limit extreme earthquake displacement; 
2. Isolated structures with a nonlinear structure (above the isolation system) including, but not 
limited to, structures designed for forces that are less than those specified by the Provisions for 
essentially-elastic design; and 
3. Isolated structures located on Class F site (that is, very soft soil). 
Lower-bound limits on isolation system design displacements and structural-design forces are specified 
by the Provisions in Sec. 13.4 as a percentage of the values prescribed by the equivalent-lateral-force 
design formulas, even when dynamic analysis is used as the basis for design. These lower-bound limits 
on key design parameters ensure consistency in the design of isolated structures and serve as a safety 
net against gross under-design. Table C13.2-2 provides a summary of the lower-bound limits on 
dynamic analysis specified by the Provisions. 
13.2.4.3 Variations in material properties: For analysis, the mechanical properties of seismic 
isolators are generally based on values provided by isolator manufacturers. The properties are evaluated 
by prototype testing, which often occurs shortly after the isolators have been manufactured, and checked 
with respect to the values assumed for design. Unlike conventional materials whose properties do not 
vary substantially with time, seismic isolators are composed of materials whose properties will generally 
vary with time. Because (a) mechanical properties can vary over the life span of a building, and (b) the 
testing protocol of Section 13.6 cannot account for the effects of aging, contamination, scragging 
(temporary degradation of mechanical properties with repeated cycling), temperature, velocity effects, 
and wear, the engineer-of-record must account for these effects by explicit analysis. One strategy for 
accommodating these effects makes use of property modification factors, which was introduced by 
Constantinou et al. (1999) in the AASHTO Guide Specification for Seismic Isolation Design 
(AASHTO, 1999). Constantinou et al. (1999) also provides information on variations in material 
properties for sliding isolation systems. Thompson et al. (2000) and Morgan et al. (2001) provide 
information on variations in material properties for elastomeric bearings. 

Seismically Isolated Structure Design Requirements 
277 
Table C13.2-2 Lower-Bound Limits on Dynamic Procedures Specified in Relation to 
ELF Procedure Requirements 
Dynamic Procedure 
Design Parameter 
ELF Procedure 
Response 
Spectrum 
Response 
History 
Design displacement  DD 
DD = (g/4p2)(SD1TD/BD) 
B 
B 
Total design displacement - DT 
DT = 1.1D 
= 0.9DT 
= 0.9DT 
Maximum displacement  DM 
DM = (g/4p2)(SM1TM/BM) 
B 
B 
Total maximum displacement - DTM 
DTM = 1.1DM 
= 0.8DTM 
= 0.8DTM 
Design shear  Vb 
(at or below the isolation system) 
Vb = kDmaxDD 
= 0.9Vb 
= 0.9Vb 
Design shear  Vs 
(regular superstructure) 
Vs = kDmaxDD/RI 
= 0.8Vs 
= 0.6Vs 
Design shear  Vs 
(irregular superstructure) 
Vs = kDmaxDDRI 
= 1.0Vs 
= 0.8Vs 
Drift (calculated using RI for Cd) 
0.015hsx 
0.015hsx 
0.020hsx 
13.2.5 Isolation system 
13.2.5.1 Environmental conditions. Environmental conditions that may adversely affect isolation 
system performance should be thoroughly investigated. Significant research has been conducted on the 
effects of temperature, aging, etc., on isolation systems since the 1970s in Europe, New Zealand, and the 
United States. 
13.2.5.2 Wind forces. Lateral displacement over the depth of the isolator zone resulting from wind 
loads should be limited to a value similar to that required for other story heights. 
13.2.5.3 Fire resistance. In the event of a fire, the isolation system should be capable of supporting the 
weight of the building, as required for other vertical-load-supporting elements of the structure, but may 
have diminished functionality for lateral (earthquake) load. 
13.2.5.4 Lateral-restoring force. The isolation system should be configured with a lateral-restoring 
force sufficient to avoid significant residual displacement as a result of an earthquake, such that the 
isolated structure will not have a stability problem so as to be in a condition to survive aftershocks and 
future earthquakes. 
13.2.5.5 Displacement restraint. The use of a displacement restraint is not encouraged by the 
Provisions. Should a displacement restraint system be implemented, explicit analysis of the isolated 
structure for maximum considered earthquake is required to account for the effects of engaging the 
displacement restraint. 
13.2.5.6 Vertical-load stability. The vertical loads to be used in checking the stability of any given 
isolator should be calculated using bounding values of dead load and live load and the peak earthquake 
demand of the maximum considered earthquake. Since earthquake loads are reversible in nature, peak 
earthquake load should be combined with bounding values of dead and live load in a manner which 
produces both the maximum downward force and the maximum upward force on any isolator. Stability 
of each isolator should be verified for these two extreme values of vertical load at peak maximum 
considered earthquake displacement of the isolation system. 
2003 Commentary, Chapter 13 
278 
13.2.5.7 Overturning. The intent of this requirement is to prevent both global structural overturning 
and overstress of elements due to local uplift. Uplift in a braced frame or shear wall is acceptable so 
long as the isolation system does not disengage from its horizontal-resisting connection detail. The 
connection details used in some isolation systems are such that tension is not permitted on the system. 
If the tension capacity of an isolation system is to be utilized to resist uplift forces, then component tests 
should be performed to demonstrate the adequacy of the system to resist tension forces at the design 
displacement. 
13.2.5.8 Inspection and replacement. Although most isolation systems will not need to be replaced 
after an earthquake, it is good practice to provide for inspection and replacement. After an earthquake, 
the building should be inspected and any damaged elements should be replaced or repaired. It is 
advised that periodic inspections be made of the isolation system. 
13.2.5.9 Quality control. A test and inspection program is necessary for both fabrication and 
installation of the isolation system. Because base isolation is a developing technology, it may be 
difficult to reference standards for testing and inspection. Reference can be made to standards for some 
materials such as elastomeric bearings (ASTM D 4014). Similar standards are required for other 
isolation systems. Special inspection procedures and load testing to verify manufacturing quality should 
be developed for each project. The requirements will vary with the type of isolation system used. 
13.2.6 Structural system 
13.2.6.1 Horizontal distribution of force 
13.2.6.2 Building separations. A minimum separation between the isolated structure and a rigid 
obstruction is required to allow free movement of the superstructure in all lateral directions during an 
earthquake. Provision should be made for lateral motion greater than the design displacement, since the 
exact upper limit of displacement cannot be precisely determined. 
13.2.7 Elements of structures and nonstructural components. To accommodate the differential 
movement between the isolated building and the ground, provision for flexible utility connections 
should be made. In addition, rigid structures crossing the interface (such as stairs, elevator shafts and 
walls) should have details to accommodate differential motion at the isolator level without sustaining 
damage sufficient to threaten life safety. 
13.3 EQUIVALENT LATERAL FORCE PROCEDURE 
13.3.2 Minimum lateral displacements. The lateral displacement given by Eq. 13.3-1 approximates 
peak design earthquake displacement of a single-degree-of-freedom, linear-elastic system of period, TD, 
and equivalent viscous damping, D, and the lateral displacement given by Eq. 13.3-3 approximates 
peak maximum considered earthquake displacement of a single-degree-of-freedom, linear-elastic system 
of period, TM, and equivalent viscous damping, DM. 
Equation 13.3-1 is an estimate of peak displacement in the isolation system for the design earthquake. 
In this equation, the spectral acceleration term, SD1, is the same as that required for design of a 
conventional fixed-base structure of period, TD. A damping term, BD, is used to decrease (or increase) 
the computed displacement when the equivalent damping coefficient of the isolation system is greater 
(or smaller) than 5 percent of critical damping. Values of coefficient BD (or BM for the maximum 
considered earthquake) are given in Table 13.3-1 for different values of isolation system damping, D 
(or M). 
A comparison of values obtained from Eq. 13.3-1 and those obtained from nonlinear time-history 
analyses are given in Kircher et al. (1988) and Constantinou et al. (1993). 
Consideration should be given to possible differences in the properties of the isolation system used for 
design and the properties of isolation system actually installed in the building. Similarly, consideration 
should be given to possible changes in isolation system properties due to different design conditions or 
load combinations. If the true deformational characteristics of the isolation system are not stable or vary 
Seismically Isolated Structure Design Requirements 
279 
with the nature of the load (being rate-, amplitude-, or time-dependent), the design displacements should 
be based on deformational characteristics of the isolation system that give the largest possible deflection 
(kDmin), the design forces should be based on deformational characteristics of the isolation system that 
give the largest possible force (kDmax), and the damping level used to determine design displacements 
and forces should be based on deformational characteristics of the isolation system that represent the 
minimum amount of energy dissipated during cyclic response at the design level. 
The configuration of the isolation system for a seismically isolated building or structure should be 
selected in such a way as to minimize any eccentricity between the center of mass of the superstructure 
and the center of rigidity of the isolation system. In this way, the effect of torsion on the displacement 
of isolation elements will be reduced. As for conventional structures, allowance for accidental 
eccentricity in both horizontal directions must be considered. Figure C13.3-1 defines the terminology 
used in the Provisions. Equation 13.3-5 (or Eq. 13.3-6 for the maximum considered earthquake) 
provides a simplified formulae for estimating the response due to torsion in lieu of a more refined 
analysis. The additional component of displacement due to torsion increases the design displacement at 
the corner of the structure by about 15 percent (for a perfectly square building in plan) to about 30 
percent (for a very long, rectangular building) if the eccentricity is 5 percent of the maximum plan 
dimension. Such additional displacement, due to torsion, is appropriate for buildings with an isolation 
system whose stiffness is uniformly distributed in plan. Isolation systems that have stiffness 
concentrated toward the perimeter of the building or certain sliding systems that minimize the effects of 
mass eccentricity will have reduced displacements due to torsion. The Provisions permits values of DT 
as small as 1.1DD, with proper justification. 
Figure C13.3-1 Displacement terminology. 
2003 Commentary, Chapter 13 
280 
13.3.3 Minimum lateral forces. Figure C13.3-2 defines the terminology below and above the 
isolation system. Equation 13.3-7 gives peak seismic shear on all structural components at or below the 
seismic interface without reduction for ductile response. Equation 13.3-8 specifies the peak seismic 
shear for design of structural systems above the seismic interface. For structures that have appreciable 
inelastic-deformation capability, this equation includes an effective reduction factor of up to 2 for 
response beyond the strength-design level. 
The basis for the reduction factor is that the design of the structural system is based on strength-design 
procedures. A factor of at least 2 is assumed to exist between the design-force level and the true-yield 
level of the structural system. An investigation of 10 specific buildings indicated that this factor varied 
between 2 and 5 (ATC, 1982). Thus, a reduction factor of 2 is appropriate to ensure that the structural 
system remains essentially elastic for the design earthquake . 
In Sec. 13.3.3.2, the limitations given on VS ensure that there is at least a factor of 1.5 between the 
nominal yield level of the superstructure and (1) the yield level of the isolation system, (2) the ultimate 
capacity of a sacrificial wind-restraint system which is intended to fail and release the superstructure 
during significant lateral load, or (3) the break-away friction level of a sliding system. 
These limitations are essential to ensure that the superstructure will not yield prematurely before the 
isolation system has been activated and significantly displaced. 
The design shear force, VS, specified by the requirements of this section ensures that the structural 
system of an isolated building will be subjected to significantly lower inelastic demands than a 
conventionally designed structure. Further reduction in VS, such that the inelastic demand on a 
seismically isolated structure would be the same as the inelastic demand on a conventionally designed 
structure, was not considered during development of these requirements but may be considered in the 
future. 
If the level of performance of the isolated structure is desired to be greater than that implicit in these 
requirements, then the denominator of Eq. 13.3-8 may be reduced. Decreasing the denominator of Eq. 
13.3-8 will lessen or eliminate inelastic response of the superstructure for the design-basis event. 
13.3.4 Vertical distribution of forces. Equation 13.3-9 describes the vertical distribution of lateral 
force based on an assumed triangular distribution of seismic acceleration over the height of the structure 
Figure C13.3-2 Isolation system 
terminology
Seismically Isolated Structure Design Requirements 
281 
above the isolation interface. Constantinou et al. (1993) provides a good summary of recent work which 
demonstrates that this vertical distribution of force will always provide a conservative estimate of the 
distributions obtained from more detailed, nonlinear analysis studies. 
13.3.5 Drift limits. The maximum story drift permitted for design of isolated structures varies 
depending on the method of analysis used, as summarized in Table C13.3-1. For comparison, the drift 
limits prescribed by the Provisions for fixed-base structures also are summarized in Table C13.3-1. 
Table C13.3-1 Comparison of Drift Limits for Fixed-Base and Isolated Structures 
Structure 
Seismic Use Group 
Fixed-Base 
Isolated 
I 
0.025hsx/(Cd/R) 
0.015hsx 
II 
0.020hsx/(Cd/R) 
0.015hsx 
Buildings (other than 
masonry) four stories or 
less in height with 
component drift design 
III 
0.015hsx/(Cd/R) 
0.015hsx 
I 
0.020hsx/(Cd/R) 
0.015hsx 
II 
0.015hsx/(Cd/R) 
0.015hsx 
Other (non-masonry) 
buildings 
III 
0.010hsx/(Cd/R) 
0.015hsx 
Drift limits in Table C13.3-1 are divided by Cd/R for fixed-base structures since displacements 
calculated for lateral loads reduced by R are factored by Cd before checking drift. The Cd term is used 
throughout the Provisions for fixed-base structures to approximate the ratio of actual earthquake 
response to response calculated for reduced forces. Generally, Cd is 1/2 to 4/5 the value of R. For 
isolated structures, the RI factor is used both to reduce lateral loads and to increase displacements 
(calculated for reduced lateral loads) before checking drift. Equivalency would be obtained if the drift 
limits for both fixed-base and isolated structures were based on their respective R factors. It may be 
noted that the drift limits for isolated structures are generally more conservative than those for 
conventional, fixed-base structures, even when fixed-base structures are designed as Seismic Use Group 
III buildings. 
13.4 DYNAMIC PROCEDURES 
This section specifies the requirements and limits for dynamic procedures. The design displacement 
and force limits on response spectrum and response history procedures are given in Table C13.2-1. 
A more-detailed or refined study can be performed in accordance with the analysis procedures described 
in this section. The intent of this section is to provide procedures which are compatible with the 
minimum requirements of Sec. 13.3. Reasons for performing a more refined study include: 
1. The importance of the building. 
2. The need to analyze possible structure/isolation-system interaction when the fixed-base period of 
the building is greater than one third of the isolated period. 
3. The need to explicitly model the deformational characteristics of the lateral-force-resisting system 
when the structure above the isolation system is irregular. 
4. The desirability of using site-specific ground-motion data, especially for soft soil types (Site Class 
F) or for structures located where S1 is greater than 0.60. 
5. The desirability of explicitly modeling the deformational characteristics of the base-isolation 
system. This is especially important for systems that have damping characteristics that are 
amplitude-dependent, rather than velocity-dependent, since it is difficult to determine an appropriate 
value of equivalent viscous damping for these systems. 
2003 Commentary, Chapter 13 
282 
Sec. 13.2.4 of this commentary discusses other conditions which require use of the response history 
procedure. 
When response history analysis is used as the basis for design, the design displacement of the isolation 
system and design forces in elements of the structure above are to be based on the maximum of the 
results of not less than three separate analyses, each using a different pair of horizontal time histories. 
Each pair of horizontal time histories should: 
1. Be of a duration consistent with the design earthquake or the maximum considered earthquake, 
2. Incorporate near-field phenomena, as appropriate, and 
3. Have response spectra for which the square-root-of-the-sum-of-the-squares combination of the two 
horizontal components equals or exceeds 1.3 times the target spectrum at each spectral ordinate. 
The average value of seven time histories is a standard required by the nuclear industry and is 
considered appropriate for nonlinear response history analysis of seismically isolated structures. 
13.5 DESIGN REVIEW 
Review of the design and analysis of the isolation system and design review of the isolator testing 
program is mandated by the Provisions for two key reasons: 
1. The consequences of isolator failure could be catastrophic. 
2. Isolator design and fabrication technology is evolving rapidly and may be based on technologies 
unfamiliar to many design professionals. 
The Provisions requires review to be performed by a team of registered design professionals that are 
independent of the design team and other project contractors. The review team should include 
individuals with special expertise in one or more aspects of the design, analysis, and implementation of 
seismic isolation systems. 
The review team should be formed prior to the development of design criteria (including site-specific 
ground shaking criteria) and isolation system design options. Further, the review team should have full 
access to all pertinent information and the cooperation of the design team and regulatory agencies 
involved with the project. 
13.6 TESTING 
The design displacements and forces developed from the Provisions are predicated on the basis that the 
deformational characteristics of the base isolation system have been previously defined by a 
comprehensive set of tests. If a comprehensive amount of test data are not available on a system, major 
design alterations in the building may be necessary after the tests are complete. This would result from 
variations in the isolation-system properties assumed for design and those obtained by test. Therefore, it 
is advisable that prototype systems be tested during the early phases of design, if sufficient test data is 
not available on an isolation system. 
Typical force-deflection (or hysteresis) loops are shown in Figure C13.6-1; also included are the 
definitions of values used in Sec. 13.6.2. 
Seismically Isolated Structure Design Requirements 
283 
The required sequence of tests will verify experimentally: 
1. The assumed stiffness and capacity of the wind-restraining mechanism; 
2. The variation in the isolator's deformational characteristics with amplitude (and with vertical load, if 
it is a vertical load-carrying member); 
3. The variation in the isolator's deformational characteristics for a realistic number of cycles of 
loading at the design displacement; and 
4. The ability of the system to carry its maximum and minimum vertical loads at the maximum 
displacement. 
Force-deflection tests are not required if similarly sized components have been tested previously using 
the specified sequence of tests. 
Variations in effective stiffness greater than 15 percent over 3 cycles of loading at a given amplitude, or 
greater than 20 percent over the larger number of cycles at the design displacement, would be cause for 
rejection. The variations in the vertical loads required for tests of isolators which carry vertical, as well 
as lateral, load are necessary to determine possible variations in the system properties with variations in 
overturning force. The appropriate dead loads and overturning forces for the tests are defined as the 
average loads on a given type and size of isolator for determining design properties and are the absolute 
maximum and minimum loads for the stability tests. 
13.6.4 Design properties of the isolation system 
13.6.4.1 Maximum and minimum effective stiffness. The effective stiffness is determined from the 
hysteresis loops shown in Figure C13.6-1). Stiffness may vary considerably as the test amplitude 
Figure C13.6-1 The effect of stiffness on an isolation bearing. 
2003 Commentary, Chapter 13 
284 
increases but should be reasonably stable (within 15 percent) for more than 3 cycles at a given 
amplitude. 
The intent of these requirements is to ensure that the deformational properties used in design result in 
the maximum design forces and displacements. For determining design displacement, this means using 
the lowest damping and effective-stiffness values. For determining design forces, this means using the 
lowest damping value and the greatest stiffness value. 
13.6.4.2 Effective damping. The determination of equivalent viscous damping is reasonably reliable 
for systems whose damping characteristics are velocity dependent. For systems that have amplitudedependent, 
energy-dissipating mechanisms, significant problems arise in determining an equivalent 
viscous-damping value. Since it is difficult to relate velocity and amplitude-dependent phenomena, it is 
recommended that when the equivalent-viscous damping assumed for the design of amplitudedependent, 
energy-dissipating mechanisms (such as pure-sliding systems) is greater than 30 percent, 
then the design-basis force and displacement should be determined using the response history 
procedure, as discussed in Commentary Sec. 13.2.4. 
REFERENCES 
ATC. 1982. An Investigation of the Correlation Between Earthquake Ground Motion and Building 
Performance, ATC Report 10. Redwood City, California: Applied Technology Council, 
ATC. 1986. Proceedings of a Seminar and Workshop on Base Isolation and Passive Energy 
Dissipation, ATC Report 17. Redwood City, California: Applied Technology Council. 
ATC. 1993. Proceedings of Seminar on Seismic Isolation, Passive Energy Dissipation, and Active 
Control, ATC 17-1.,Redwood City, California: Applied Technology Council. 
ATC. 2002. Seminar on Response Modification Technologies for Performance-Based Seismic Design, 
ATC 17-2. Redwood City, California: Applied Technology Council. 
AASHTO. 1999. Guide Specification for Seismic Isolation Design, Washington, D.C: American 
Association of State Highway and Transportation Officials. 
ASCE. 2003. Minimum Design Loads for Buildings and Other Structures, ASCE-7-02, Reston, VA: 
American Society of Civil Engineers. 
ASCE. 1989, 1991, 1993, and 1995. Seismic Engineering: Research and Practice. New York City: 
American Society of Civil Engineers. 
Constantinou, M. C., C. W. Winters, and D. Theodossiou. 1993. Evaluation of SEAOC and UBC 
analysis procedures, Part 2: Flexible superstructure, in Proceedings of a Seminar on Seismic Isolation, 
Passive Energy Dissipation and Active Control, ATC Report 17-1. Redwood City, California: ATC. 
Constantinou, M.C., P. Tsopelas, A. Kasalanati, and E. D. Wolff. 1999. Property modification factors 
for seismic isolation bearings, Report No. MCEER-99-0012, Multidisciplinary Center for Earthquake 
Engineering Research, Buffalo, NY. 
Earthquake Engineering Research Institute. 1990. Seismic isolation: from idea to reality, Earthquake 
Spectra Journal 6:2. 
Hussain, S.M.. 1998. Testing and quality control issues for seismic isolation systems-an overview, in 
Proceedings of of the Structural Engineers World Congress, Paper T163-4, San Francisco, CA. 
International Code Council, 2003. International Building Code, Falls Creek, VA 
Kircher, C. A., B. Lashkari, R. L. Mayes, and T. E. Kelly. 1988. Evaluation of nonlinear response in 
seismically isolated buildings, in Proceedings of a Symposium on Seismic, Shock and Vibration 
Isolation, ASTHMA PVP Conference. 
Seismically Isolated Structure Design Requirements 
285 
Lashkari, B., and C. A. Kircher. 1993. Evaluation of SEAOC & UBC analysis procedures, part 1: stiff 
superstructure, in Proceedings of a Seminar on Seismic Isolation, Passive Energy Dissipation and 
Active Control, ATC Report 17-1. Redwood City, California: ATC. 
Morgan, T., A. S. Whittaker, and A. C. Thompson. 2001. Cyclic behavior of high-damping rubber 
bearings, Proceedings, Fifth World Congress on Joints, Bearings and Seismic Systems for Concrete 
Structures, American Concrete Institute, Rome, Italy. 
Skinner, R. I., W. H. Robinson, and G. H. McVerry. 1993. An Introduction to Seismic Isolation. 
Sussex, England: Wiley and Sons. 
Thompson, A., A. S. Whittaker, G. L. Fenves, and S. A. Mahin. 2000. Property modification factors for 
elastomeric bearings, Proceedings, Twelfth World Conference on Earthquake Engineering, Auckland, 
New Zealand. 

2003 Commentary, Chapter 13 
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 287 
Chapter 14 Commentary 
NONBUILDING STRUCTURE DESIGN REQUIREMENTS 
14.1 GENERAL 
14.1.1 Scope. Requirements concerning nonbuilding structures were originally added to the 1994 
Provisions by the 1991-94 Provisions Update Committee (PUC) at the request of the BSSC Board of 
Direction to provide building officials with needed guidance. In recognition of the complexity, nuances, 
and importance of nonbuilding structures, the BSSC Board established 1994-97 PUC Technical 
Subcommittee 13 (TS13), Nonbuilding Structures, in 1995. The duties of TS13 were to review the 1994 
Provisions and Commentary and recommend changes for the 1997 Edition. The subcommittee comprised 
individuals possessing considerable expertise concerning various specialized nonbuilding structures and 
representing a wide variety of industries concerned with nonbuilding structures. 
Building codes traditionally have been perceived as minimum standards of care for the design of 
nonbuilding structures and building code compliance of these structures is required by building officials 
in many jurisdictions. However, requirements in the industry standards are often at odds with building 
code requirements. In some cases, the industry standards need to be altered while in other cases the 
building codes need to be modified. Registered design professionals are not always aware of the 
numerous accepted standards within an industry and may not know whether the accepted standards are 
adequate. It is hoped that Chapter 14 of the Provisions appropriately bridges the gap between building 
codes and existing industry standards. 
One of the goals of TS13 was to review and list appropriate industry standards to serve as a resource. 
These standards had to be included in the appendix. The subcommittee also has attempted to provide an 
appropriate link so that the accepted industry standards can be used with the seismic ground motions 
established in the Provisions. It should be noted that some nonbuilding structures are very similar to a 
building and can be designed employing sections of the Provisions directly whereas other nonbuilding 
structures require special analysis unique to the particular type of nonbuilding structure. 
The ultimate goal of TS13 was to provide guidance to develop requirements consistent with the intent of 
the Provisions while allowing the use of accepted industry standards. Some of the referenced standards 
are consensus documents while others are not. 
One good example of the dilemma posed by the conflicts between the Provisions and accepted design 
practice for nonbuilding structures involves steel multilegged water towers. Historically, such towers 
have performed well when properly designed in accordance with American Water Works Association 
(AWWA) standards, but these standards differ from the Provisions that tension-only rods are required and 
the connection forces are not amplified. However, industry practice requires upset rods that are preloaded 
at the time of installation, and the towers tend to perform well in earthquake areas. 
In an effort to provide the appropriate interface between the Provisions requirements for building 
structures, nonstructural components, and nonbuilding structures; TS13 recommended that nonbuilding 
structure requirements be placed in a separate chapter. The PUC agreed with this change. The 1997 
Provisions Chapter 14 now provides registered design professionals responsible for designing 
nonbuilding structures with a single point of reference. 
Note that building structures, vehicular and railroad bridges, electric power substation equipment, 
overhead power line support structures, buried pipelines and conduits, tunnels, lifeline systems, nuclear 
power plants, and dams are excluded from the scope of the nonbuilding structure requirements. The 
excluded structures are covered by other well established design criteria (e.g., electric power substation 
equipment, power line support structures, vehicular and railroad bridges), are not under the jurisdiction of 
2 003 Commentary, Chapter 14 
288 
local building officials (e.g., nuclear power plants, and dams), or require technical considerations beyond 
the scope of the Provisions (e.g., piers and wharves, buried pipelines and conduits, tunnels, and lifeline 
systems). Since many components of lifeline systems can be designed in accordance with the Provisions, 
the following information is provided to clarify why lifeline systems are excluded from the scope of the 
Provisions. 
Seismic design for a lifeline system will typically require consideration of factors that are unique to or 
particularly important to that specific system. Seismic design requirements for lifeline systems will 
typically differ from those for buildings individual structural components for the following reasons: 
1. Physical characteristics. A building consists of structural and non-structural components within a 
single site, whereas lifeline systems consist of networks of multiple and spatially distributed linked 
components (primarily non-building structures and equipment, and possibly some buildings as well.) 
2. Stakeholders. The stakeholders in the continued operation of a building after an earthquake are a 
relatively small group of building owners, tenants, and insurers. Lifeline systems provide essential 
services to a community (e.g., electric power,, communications, transportation, natural gas, water, 
wastewater, and liquid fuel). Therefore, stakeholders in the seismic performance of such systems are 
the businesses and residents of the region served by the system, business clients/vendors outside of 
the region whose continued operation will be impacted by the conditions of the businesses/residents 
within the region, and the lifeline systems owners and insurers. 
3. Performance.. Acceptable seismic performance of a building is typically measured by whether life 
safety of building occupants has been adequately protected (in accordance with minimum building 
code design provisions.) In addition, for those relatively few buildings for which performance based 
design has been considered, acceptable seismic performance will also be measured by how well postearthquake 
functionality and return-to-service requirements of the building tenants have been met. 
The ability of a lifeline system to maintain an acceptable level of service after an earthquake will 
depend, not only on the seismic performance of its various spatially dispersed components, but also 
on the redundancy and service capacity of these components (e.g., number of lanes within roadway 
elements). To the extent that a lifeline system is comprised of redundant components of sufficient 
service capacity, it can maintain an acceptable level of service to a community even if some of the 
redundant components are damaged during the earthquake. In addition, except for certain 
transportation structures (e.g., bridges and tunnels), earthquake damage to the lifeline system 
components generally do not result in direct life-safety consequences. Therefore, acceptable seismic 
performance for a lifeline system is typically based on: (a) whether the system provides an adequate 
level of service to its users after an earthquake; (b) whether economic losses related to direct damage, 
lost revenue from an inoperable system, and liability exposure are within tolerable limits; and (c) 
whether any adverse political, legal, social, administrative, or environmental consequences are 
experienced. For these reasons, acceptable seismic performance requirements for lifeline systems are 
best established through interaction with the appropriate stakeholders, including the lifeline agency, 
its customers or users, and appropriate regulatory interests. 
The definition of what constitutes a component of a lifeline system is often complicated. Components of 
utility lifeline systems are typically identical to components that might be found in industrial or 
commercial applications. A good example of this overlap are aboveground storage tanks that are 
common in large industrial or manufacturing facilities as well as water and liquid hydrocarbon 
transportation systems. Because of this similarity, a clear definition is needed to determine when design 
in accordance with the recommended approach for lifeline systems should be give preference over 
requirements in the Provisions. Three criteria are considered for determining whether the design of a 
particular nonbuilding structure can be treated as a component of a lifeline system. 
1. Spatial distribution. As noted above, lifeline systems are typically spatially-distributed systems that 
provide services considered essential to community activities and include electric power, 
communications, water, waste-water, natural gas, liquid fuel, and transportation systems. Fixed 
facilities, such as power plants, compressor stations, metering stations, are typically treated as nodes 
 Nonbuilding Structure Design Requirements 
289 
of a lifeline system and are designed in accordance with these Provisions. 
2. Definition by legal boundary. Portions of utility lifeline systems upstream of the point defining the 
legal boundary for ownership and responsibility for maintenance and repair shall be considered as 
part of a lifeline system. The physical elements of transportation lifeline systems not excluded in the 
Provisions and owned and maintained by a transportation agency are also considered part of a lifeline 
system. 
Defining lifeline system components by a legal boundary is most appropriate for utility systems that 
deliver electric power, natural gas, electric power, wastewater and some telecommunication services. 
Existing regulatory provisions commonly specify a specific interface between the portions of these 
systems that is under the control of the service provider and the portions of the system under control 
of the building or facility owner. For electric power, natural gas, and water systems, this boundary is 
typically the customers side of the meter. The other typical boundary is the property line. Those 
components under control of the service provider can be considered as part of a lifeline system. 
It is common for the design and maintenance of physical elements of transportation lifeline systems to 
fall under the jurisdiction of a governmental or government-regulated entity. Two common examples 
include state highway departments and port authorities. In such cases, the definition of a lifeline 
system by legal boundary for these situations is defined by the jurisdiction of these agencies. 
3. Definition by expertise. Historically, the primary audience of the Provisions has been the structural 
engineering community and building code organizations seeking to modify their seismic provisions. 
As a result of this focus, the Provisions are best suited for the seismic design and performance of 
individual structures. Since most new construction for lifeline systems address adding components to 
existing systems, rational design approaches should consider the overall system performance in 
design of new components and the benefits of improved seismic performance in comparison with the 
performance of the system for other natural and other hazards, such as man-made threats. The 
geographically diverse nature of lifeline systems often requires that earthquake hazards be defined by 
one or more scenario events instead of the probabilistic ground motion hazards defined in the 
Provisions. These additional considerations often require special expertise in addition to that of the 
structural engineering profession that is dominant audience for the Provisions. 
14.1.2 References 
American Concrete Institute (ACI): 
ACI 350, Environmental Engineering Concrete Structures, 2001. 
ACI 350.3 Seismic Design of Liquid-Containing Concrete Structures, 2001. 
American Society of Civil Engineers (ASCE): 
Guidelines for Seismic Evaluation and Design of Petrochemical Facilities, 1997. 
Housner, G.W. Earthquake Pressures in Fluid Containers, California Institute of Technology, 1954. 
Miller, C. D., S. W. Meier, and W. J. Czaska, Effects of Internal Pressure on Axial Compressive Strength 
of Cylinders and Cones, Structural Stability Research Council Annual Technical Meeting, June 1997. 
Rack Manufacturers Institute: 
Specification for the Design, Testing, and Utilization of Industrial Steel Storage Racks, 1997 
(Reaffirmed 2002). 
Troitsky, M.S., Tubular Steel Structures, 1990. (Troitsky) 
Wozniak, R. S., and W. W. Mitchell, Basis of Seismic Design Provisions for Welded Steel Oil Storage 
Tanks, 1978 ProceedingsRefining Dept., Vol. 57, American Petroleum Institute, Washington, D.C., 
May 9, 1978. (Wozniak) 
2 003 Commentary, Chapter 14 
290 
14.1.2.2 Other references 
While not cited directly in the Provisions or Commentary, the user may find these other references related 
to nonbuilding structures helpful. 
ACI 307, Standard Practice for the Design and Construction of Cast-In-Place Reinforced 
Concrete Chimneys, 1995. 
ANSI K61.1 Safety Requirements for the Storage and Handling of Anhydrous Ammonia, 
American National Standards Institute, 1999. 
API 2510 Design and Construction of Liquefied Petroleum Gas Installation, American 
Petroleum Institute, 19952001. 
ASCE Petro Design of Secondary Containment in Petrochemical Facilities 
(Petrochemical Energy Committee Task Report), American Society of Civil 
Engineers, 1997. 
ASME B31.8 Gas Transmission and Distribution Piping Systems, American Society of 
Mechanical Engineers, 1995. 
ASME B96.1 Welded Aluminum-Alloy Storage Tanks, American Society of Mechanical 
Engineers, 1993. 
ASME STS-1 Steel Stacks, American Society of Mechanical Engineers, 2001. 
ASTM F 1159 Standard Practice for the Design and Manufacture of Amusement Rides and 
Devices (ASTM F 1159-97a), American Society for Testing and Materials, 1997. 
ASTM C 1298 Standard Guide for Design and Construction of Brick Liners for Industrial 
Chimneys (ASTM C 1298-95), American Society for Testing and Materials, 
1995. 
DOT 49CFR193 Liquefied Natural Gas Facilities: Federal Safety Standards (Title 49CFR Part 
193), U.S. Department of Transportation, 2000. 
NFPA 30 Flammable and Combustible Liquids Code, National Fire Protection Association, 
12000. 
NFPA 58 Storage and Handling of Liquefied Petroleum Gas, National Fire Protection 
Association, 2001. 
NFPA 59 Storage and Handling of Liquefied Petroleum Gases at Utility Gas Plants, 
National Fire Protection Association, 2001. 
NFPA 59A Production, Storage and Handling of Liquefied Natural Gas (LNG), National 
Fire Protection Association, 2001. 
NCEL R-939 Ebeling, R. M., and Morrison, E. E., The Seismic Design of Waterfront Retaining 
Structures, Naval Civil Engineering Laboratory, 1993. 
NAVFAC DM-25.1 Piers and Wharves, U.S. Naval Facilities Engineering Command, 1987. 
TM 5-809-10 Seismic Design for Buildings, U.S. Army Corps of Engineers, 1992, Chapter 13 
only. 
14.1.5 Nonbuilding structures supported by other structures. This section has been developed to 
provide an appropriate link between the requirements for nonbuilding structures and those for inclusion in 
the rest of the Provisionsespecially the requirements for architectural, mechanical, and electrical 
components. 
14.2 GENERAL DESIGN REQUIREMENTS 
14.2.1 Seismic use groups and importance factors. The Importance Factors and Seismic Use Group 
classifications assigned to nonbuilding structures vary from those assigned to building structures. 
 Nonbuilding Structure Design Requirements 
291 
Buildings are designed to protect occupants inside the structure whereas nonbuilding structures are not 
normally occupied in the same sense as buildings, but need to be designed in a special manner because 
they pose a different sort of risk in regard to public safety (that is, they may contain very hazardous 
compounds or be essential components in critical lifeline systems). For example, tanks and vessels may 
contain materials that are essential for lifeline functions following a seismic event (such as fire-fighting or 
potable water), potentially harmful or hazardous to the environment or general health of the public, 
biologically lethal or toxic, or explosive or flammable (posing a threat of consequential or secondary 
damage). 
If not covered by the authority having jurisdiction, Table 14.2-1 may be used to select the importance 
factor (I). The value shall be determined by taking the larger of the value from the approved Standard or 
the value selected from Table 14.2-1. It should be noted that a single value of importance factor may not 
apply to an entire facility. For further details, refer to ASCE Petro. The use of a secondary containment 
system, when designed in accordance with an acceptable National Standard, could be considered as an 
effective means to contain hazardous substances and thus reduce the hazard classification. 
The specific definition of material hazard and what constitutes a hazard is being developed in the 
International Building Code process. The hazards will be predicated on the quantity and type of 
hazardous material. 
The importance factor is not intended for use in making economic evaluations regarding the level of 
damage, probabilities of occurrence, or cost to repair the structure. These economic decisions should be 
made by the owner and other interested parties (insurers, financiers, etc.). Nor it is intended for use for 
purposes other than that defined in this provision. 
Examples are presented below demonstrate how this table may be applied. 
Example 1. A water storage tank used to provide pressurized potable water for a process within a 
chemical plant where the tank is located away from personnel working within the facility. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues implied in the matrix: 
$ Seismic Use Group: Neither the structure nor the contents are critical, therefore use Seismic Use 
Group I. 
$ Function: The water storage tank is neither a designated ancillary structure for post-earthquake 
recovery, nor identified as an emergency back-up facilities for a Seismic Use Group III structure, 
therefore use F-I. 
$ Hazard: The contents are not hazardous, therefore use H-I. 
$ This tank has an importance factor of 1.0. 

2 003 Commentary, Chapter 14 
292 
Example 2. A steel storage rack is located in a retail store in which the customers have direct access to 
the aisles. Merchandise is stored on the upper racks. The rack is supported by a slab on grade. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: Neither the structure nor the contents are critical, therefore use Seismic Use 
Group I. 
$ Function: The storage rack is neither used for post-earthquake recovery, nor required for emergency 
back-up, therefore use F-I. 
$ Hazard: The contents are not hazardous. However, its use could cause a substantial public hazard 
during an earthquake. Subject to the local authoritys jurisdiction it is H-II. 
$ According to Sec. 14.3.5.2 the importance factor for storage racks in occupancies open to the general 
public must be taken as 1.5. 
$ Use an importance factor of 1.5 for this structure. 
Example 3. A water tank is located within an office building complex to supply the fire sprinkler system. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: The office building is assigned to Seismic Use Group I. 
$ Function: The water tank is required to provide water for fire fighting. However since the building is 
not a Seismic Use Group III structure, the water is used neither for post-earthquake recovery, nor for 
emergency back-up, so use F-I. 
$ Hazard: The content and its use are not hazardous to the public, therefore use H-I. 
$ Use an importance factor of 1.0 for this water structure. 

 Nonbuilding Structure Design Requirements 
293 
Example 4. A petrochemical storage tank is to be constructed within a refinery tank farm near a 
populated city neighborhood. An impoundment dike is provided to control liquid spills. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: The LNG tank is assigned to Seismic Use Group III. 
$ Function: The tank is neither required to provide post-earthquake recovery nor used for emergency 
back-up for a Seismic Use Group III structure, so use F-I. 
$ Hazard: The tank contains a substantial quantity of high explosive and is near a city neighborhood. 
Despite the diking, it is considered hazardous to the public in the event of an earthquake, so use H-III. 
$ Use an importance factor of 1.5 for this structure. 
14.2.3 Design basis. The design basis for nonbuilding structures is based on either adopted references, 
approved standards, or these Provisions. It is intended that the Provisions applicable to buildings apply to 
nonbuilding structures, unless specifically noted in this Chapter. 
14.2.4 Seismic force-resisting system selection and limitations. Nonbuilding structures similar to 
buildings may be designed in accordance with either Table 4.3-1 or Table 14.2-2, including referenced 
design and detailing requirements. For convenience, Table 4.3-1 requirements are repeated in Table 14.2- 
2. 
Table 14.2-2 of the 2000 NEHRP Provisions for nonbuilding structures similar to buildings prescribed R, 
Oo
, and Cd values to be taken from Table 4.3-1, but prescribed less restrictive height limitations than 
those prescribed in Table 4.3-1. This inconsistency has been corrected. Nonbuilding structures similar to 
buildings which use the same R, Oo
, and Cd values as buildings now have the same height limits, 
restrictions and footnote exceptions as buildings. The only difference is that the footnote exceptions for 
buildings apply to metal building like systems while the exceptions for nonbuilding structures apply to 
pipe racks. In addition, selected nonbuilding structures similar to buildings have prescribed an option 
where both lower R values and less restrictive height limitations are specified. This option permits 
selected types of nonbuilding structures which have performed well in past earthquakes to be constructed 
with less restrictions in Seismic Design Categories D, E and F provided seismic detailing is used and 
design force levels are considerably higher. It should be noted that revised provisions are considerably 
more restrictive than those prescribed in Table 4.3-1. 
Nonbuilding structures not similar to buildings should be designed in accordance Table 14.2-3 
requirements, including referenced design and detailing requirements. 
Nonbuilding structures not referenced in either Table 14.2-2, Table 14.2-3, or Table 4.3-1 may be 
designed in accordance with an adopted reference, including its design and detailing requirements. 
It is not consistent with the intent of the Provisions to take design values from one table or standard and 
design and/or detailing provisions from another. 

2 003 Commentary, Chapter 14 
294 
14.2.5 Structural analysis procedure selection. Nonbuilding structures that are similar to buildings 
should be subject to the same analysis procedure limitations as building structures. 
Nonbuilding structures that are not similar to buildings should not be subject to these procedure 
limitations. However, they should be subject to any procedure limitations prescribed in specific adopted 
references. 
For nonbuilding structures supporting flexible system components, such as pipe racks, the supported 
piping and platforms are generally not regarded as rigid enough to redistribute seismic forces to the 
supporting frames. 
For nonbuilding structures supporting rigid system components, such as steam turbine generators (STGs) 
and Heat Recovery Steam Generators (HRSGs), the supported equipment, ductwork, and other 
components (depending on how they are attached to the structure) may be rigid enough to redistribute 
seismic forces to the supporting frames. Torsional effects may need to be considered in such situations. 
14.2.9 Fundamental period. The rational methods for period calculation contained in the Provisions 
were developed for building structures. If the nonbuilding structure has dynamic characteristics similar to 
those of a building, the difference in period is insignificant. If the nonbuilding structure is not similar to a 
building structure, other techniques for period calculation will be required. Some of the references for 
specific types of nonbuilding structures contain more accurate methods for period determination. 
Equations 5.2-6, 5.2-7, and 5.2-7 are not recommended because they are not relevant for the commonly 
encountered nonbuilding structures. 
14.3 NONBUILDING STRUCTURES SIMILAR TO BUILDINGS 
Nonbuilding structures exhibit behavior similar to that of building structures; however, their function and 
performance are different. Although the Provisions for buildings are used as the primary basis for design, 
this section identifies appropriate exceptions, modifications, and additions for selected nonbuilding 
structures similar to buildings. 
14.3.1 Electrical power generating facilities. Electrical power plants closely resemble building 
structures, and their performance in seismic events has been good. For reasons of mechanical 
performance, lateral drift of the structure must be limited. The lateral bracing system of choice has been 
the concentrically braced frame. The height limits on braced frames in particular can be an encumbrance 
to the design of large power generation facilities. 
14.3.3 Piers and wharves. Current industry practice recognizes the distinct differences between the two 
categories of piers and wharves described in the Provisions. The piers and wharves with public 
occupancy, described in paragraph (a) are commonly treated as the foundation for buildings or 
building-like structures, and design is performed using the Provisions. The design is likely to be under 
the jurisdiction of the local building official. 
Piers and wharves where occupancy by the general public is not a consideration, as described in 
paragraph (b), are often treated differently. In many cases, they do not fall under the jurisdiction of 
building officials, and utilize other design approaches more common to this industry. 
Economics plays a major role in the design decisions associated with these structures. These economic 
decisions may be affected not only by the wishes of the owners, but also by overlapping jurisdictional 
entities with local, regional, or state interests in commercial development. 
In the cases where the Building Officials have jurisdiction, they typically do not have experience 
analyzing pier and wharf structures. In these instances, they have come to rely on and utilize the other 
design approaches that are more common in the industry. 
Major ports and marine terminals in seismic regions of the world routinely design structures as described 
in paragraph (b). The design of these often uses a performance-based approach, with criteria and methods 
that are very different than those used for buildings, as provided in the Provisions. 
Design approaches most commonly used are generally consistent with the practices and criteria described 
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295 
in the following documents: 
 Working Group No. 34 of the Maritime Navigation Commission 
(PIANC/MarCom/WG34), 2001, Seismic Design Guidelines for Port Structures, A. A. 
Balkema, Lisse, Netherlands, 2001. 
 Ferritto, J., Dickenson, S., Priestley N., Werner, S., Taylor, C., Burke D., Seelig W., and 
Kelly, S., 1999, Seismic Criteria for California Marine Oil Terminals, Vol.1 and Vol.2, 
Technical Report TR-2103-SHR, Naval Facilities Engineering Service Center, Port 
Hueneme, CA. 
 Priestley, N.J.N., Frieder Siebel, Gian Michele Calvi, Seismic Design and Retrofit of 
Bridges, 1996, New York. 
 Seismic Guidelines for Ports, by the Ports Committee of the Technical Council on 
Lifeline Earthquake Engineering, ASCE, edited by Stuart D. Werner, Monograph No. 12, 
March 1998, published by ASCE, Reston, VA. 
 Marine Oil Terminal Engineering and Maintenance Standards, California State Lands 
Commission, Marine Facilities Division, May 2002. 
These alternative approaches have been developed over a period of many years by working groups within 
the industry, and consider the historical experience and performance characteristics of these structures 
that are very different than building structures. 
The main emphasis of the performance-based design approach is to provide criteria and methods that 
depend on the economic importance of a facility. Adherence to the performance criteria in the documents 
listed above is expected to provide as least as much inherent life-safety, and likely much more, than for 
buildings designed using the Provisions. However, the philosophy of these criteria is not to provide 
uniform margins of collapse for all structures. Among the reasons for the higher inherent level of lifesafety 
for these structures are the following: 
 These structures have relatively infrequent occupancy, with few working personnel and very 
low density of personnel. Most of these structures consist primarily of open area, with no 
enclosed building structures which can collapse onto personnel. Small control buildings on 
marine oil terminals or similar secondary structures are commonly designed in accordance 
with the local building code. 
 These pier or wharf structures are typically constructed of reinforced concrete, prestressed 
concrete, and/or steel and are highly redundant due to the large number of piles supporting a 
single wharf deck unit. Tests done for the Port of Los Angeles at the University of California 
at San Diego have shown that very high ductilities (10 or more) can be achieved in the design 
of these structures using practices currently used in California ports. 
 Container cranes, loading arms, and other major structures or equipment on the piers or 
wharves are specifically designed not to collapse in an earthquake. Typically, additional 
piles and structural members are incorporated into the wharf or pier specifically to support 
that item. 
 Experience has shown that seismic failure of wharf structures in zones of strong seismicity 
is indicated not by collapse, but by economically unrepairable deformations of the piles. The 
wharf deck generally remains level or slightly tilting but shifted out of position. Complete 
failure that could cause life-safety concerns has not been known to ever occur historically due 
to earthquake loading. 

2 003 Commentary, Chapter 14 
296 
 The performance-based criteria of the listed documents include repairability of the structure. 
This service level is much more stringent than collapse prevention and would provide a 
greater margin for life-safety. 
 Lateral load design of these structures is often governed by other marine loading conditions, 
such as mooring or berthing. 
14.3.4 Pipe racks. Free standing pipe racks supported at or below grade with framing systems that are 
similar in configuration to building systems should be designed to satisfy the force requirements of 
Sec. 5.2. Single column pipe racks that resist lateral loads should be designed as inverted pendulums. 
See ASCE Petro. 
14.3.5 Steel storage racks. This section is intended to assure comparable results from the use of the 
RMI Specification, the NEHRP Provisions, and the IBC code approaches to rack structural design. 
For many years the RMI has been working with the various committees of the model code organizations 
and with the Building Seismic Safety Council and its Technical Subcommittees to create seismic design 
provisions particularly applicable to steel storage rack structures. The 1997 RMI Specification is seen to 
be in concert with the needs, provisions, and design intent of the building codes and those who use and 
promulgate them, as well as those who engineer, manufacture, install, operate, use, and maintain rack 
structures. The RMI Specification, now including detailed seismic provisions, is essentially selfsufficient. 
The changes proposed here are compatible and coordinated with those in the 2000 International Building 
Code. 
14.3.5.2 Importance factor. Until recently, storage racks were primarily installed in low-occupancy 
warehouses. With the recent proliferation of warehouse-type retail stores, it has been judged necessary to 
address the relatively greater seismic risk that storage racks may pose to the general public, compared to 
more conventional retail environments. Under normal operating conditions, retail stores have a far higher 
occupancy load than an ordinary warehouse of a reasonable size. Failure of a storage rack system in the 
retail environment is much more likely to cause personal injury than a similar failure in a storage 
warehouse. Therefore, to provide an appropriate level of additional safety in areas open to the public, 
Sec 14.3.5.2 now requires that storage racks in occupancies open to the general public be designed with 
an importance factor equal to 1.50. Storage rack contents, while beyond the scope of the Provisions, pose 
a potentially serious threat to life should they fall from the shelves in an earthquake. Restraints should be 
provided to prevent the contents of rack shelving open to the general public from falling in strong ground 
shaking. 
14.4 NONBUILDING STRUCTURES NOT SIMILAR TO BUILDINGS 
Nonbuilding structures not similar to buildings exhibit behavior markedly different from that of building 
structures. Most of these types of structures have adopted references that address their unique structural 
performance and behavior. The ground motion in the Provisions requires appropriate translation to allow 
use with industry standards. Such translation is provided in this section. 
14.4.2 Earth retaining structures. In order to properly develop and implement methodologies for the 
design of earth retaining structures, it is essential to know and understand the nature of the applied loads. 
Concerns have been raised concerning the design of nonyielding walls and yielding walls for bending, 
overturning, sliding, etc., taking into account the varying soil types, importance, and site seismicity. See 
Sec. 7.5.1 in the Commentary. 
14.4.3 Stacks and chimneys. The design of stacks and chimneys to resist natural hazards is generally 
governed by wind design considerations. The exceptions to this general rule involve locations with high 
seismicity, stacks and chimneys with large elevated masses, and stacks and chimneys with unusual 
geometries. It is prudent to evaluate the effect of seismic loads in all but those areas with the lowest 
seismicity. Although not specifically required, it is recommended that the special seismic details required 
elsewhere in the Provisions be evaluated for applicability to stacks and chimneys. 
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297 
Guyed steel stacks and chimneys are generally light weight. As a result, the design loads due to natural 
hazards are generally governed by wind. On occasion, large flares or other elevated masses located near 
the top may require an in-depth seismic analysis. Although Chapter 6 of Troitsky does not specifically 
address seismic loading, it remains an applicable methodology for resolution of seismic forces that are 
defined in these Provisions. 
14.4.7 Tanks and vessels. Methods of seismic design of tanks, currently adopted by a number of 
industry standards, have evolved from earlier analytical work by Jacobsen, Housner, Veletsos, Haroun, 
and others. The procedures used to design flat bottom storage tanks and liquid containers is based on the 
work of Housner, Wozniak, and Mitchell. The standards for tanks and vessels have specific requirements 
to safeguard against catastrophic failure of the primary structure based on observed behavior in seismic 
events since the 1930s. Other methods of analysis using flexible shell models have been proposed but are 
presently beyond the scope of these Provisions. 
These methods entail three fundamental steps: 
1. The dynamic modeling of the structure and its contents. When a liquid-filled tank is subjected to 
ground acceleration, the lower portion of the contained liquid, identified as the impulsive component 
of mass WI, acts as if it were a solid mass rigidly attached to the tank wall. As this mass accelerates, it 
exerts a horizontal force, PI, against the wall that is directly proportional to the maximum acceleration 
of the tank base. This force is superimposed on the inertia force of the accelerating wall itself, Pw. 
Under the influence of the same ground acceleration, the upper portion of the contained liquid 
responds as if it were a solid mass flexibly attached to the tank wall. This portion, which oscillates at 
its own natural frequency, is identified as the convective component Wc, and exerts a force Pc on the 
wall. The convective component oscillations are characterized by the phenomenon of sloshing 
whereby the liquid surface rises above the static level on one side of the tank, and drops below that 
level on the other. 
2. The determination of the frequency of vibration, wI, of the tank structure and the impulsive 
component; and the natural frequency of oscillation (sloshing), wc, of the convective component. 
3. The selection of the design response spectrum. The response spectrum may be site-specific or it may 
be constructed deterministically on the basis of seismic coefficients given in national codes and 
standards. Once the design response spectrum is constructed, the spectral accelerations 
corresponding to wI and wc are obtained and are used to calculate the dynamic forces PI, Pw, and Pc. 
Detailed guidelines for the seismic design of circular tanks, incorporating these concepts to varying 
degrees, have been the province of at least four industry standards: AWWA D100 for welded steel tanks 
(since 1964); API 650 for petroleum storage tanks; AWWA D110 for prestressed, wire-wrapped tanks 
(since 1986); and AWWA D115 for prestressed concrete tanks stressed with tendons (since 1995). In 
addition, API 650 and API 620 contain provisions for petroleum, petrochemical, and cryogenic storage 
tanks. The detail and rigor of analysis employed by these standards have evolved from a semi-static 
approach in the early editions to a more rigorous approach at the present, reflecting the need to factor in 
the dynamic properties of these structures. 
The requirements in Sec 14.4.7 are intended to link the latest procedures for determining design level 
seismic loads with the allowable stress design procedures based on the methods in these Provisions. 
These requirements, which in many cases identify specific substitutions to be made in the design 
equations of the national standards, will assist users of the Provisions in making consistent 
interpretations. 
ACI has published a document, ACI 350.3-01 titled Seismic Design of Liquid-Containing Concrete 
Structures. This document, which covers all types of concrete tanks (prestressed and non-prestressed, 
circular and rectilinear), has provisions made consistent with the seismic guidelines of the 2000 Provisions. 
This ACI document serves as both a practical how-to loading reference and a guide to supplement 
application of Chapter 21 Special Provisions for Seismic Design of ACI 318. 
2 003 Commentary, Chapter 14 
298 
14.4.7.1 Design basis. Two important tasks of TS 13 were (a) to partially expand the coverage of 
nonbuilding structures in the Provisions; and (b) to provide comprehensive cross-references to all the 
applicable industry standards. It is hoped that this endeavor will bring about a standardization and 
consistency of design practices for the benefit of both the practicing engineer and the public at large. 
In the case of the seismic design of nonbuilding structures, standardization requires adjustments to 
industry standards to minimize existing inconsistencies among them. However, the standardization 
process should recognize that structures designed and built over the years in accordance with industry 
standards have performed well in earthquakes of varying severity. 
Of the inconsistencies among industry standards, the ones most important to seismic design relate to the 
base shear equation. The traditional base shear takes the following form: 
w 
V ZISCW 
R 
= 
An examination of those terms as used in the different references reveals the following: 
- ZS: The seismic zone coefficient, Z, has been rather consistent among all the standards by virtue of 
the fact that it has traditionally been obtained from the seismic zone designations and maps in the 
model building codes. 
On the other hand, the soil profile coefficient, S, does vary from one standard to another. In some 
standards these two terms are combined. 
S I: The importance factor, I, has also varied from one standard to another, but this variation is 
unavoidable and understandable owing to the multitude of uses and degrees of importance of liquidcontaining 
structures. 
S C: The coefficient C represents the dynamic amplification factor that defines the shape of the design 
response spectrum for any given maximum ground acceleration. Since coefficient C is primarily a 
function of the frequency of vibration, inconsistencies in its derivation from one standard to another 
stem from at least two sources: differences in the equations for the determination of the natural 
frequency of vibration, and differences in the equation for the coefficient itself. (For example, for the 
shell/impulsive liquid component of lateral force, the steel tank standards use a constant design 
spectral acceleration (namely, a constant C) that is independent of the impulsive period T.) In 
addition, the value of C will vary depending on the damping ratio assumed for the vibrating structure 
(usually between 2 percent and 7 percent of critical). 
Where a site-specific response spectrum is available, calculation of the coefficient C is not necessary 
B except in the case of the convective component (coefficient Cc) which is assumed to oscillate with 
0.5 percent of critical damping, and whose period of oscillation is usually high (greater than 2.5 sec). 
Since site-specific spectra are usually constructed for high damping values (3 percent to 7 percent of 
critical); and since the site-specific spectral profile may not be well-defined in the high-period range, 
an equation for Cc applicable to a 0.5 percent damping ratio is necessary in order to calculate the 
convective component of the seismic force. 
S Rw: The response modification factor, Rw, is perhaps the most difficult to quantify, for a number of 
reasons. While Rw is a compound coefficient that is supposed to reflect the ductility, energydissipating 
capacity, and redundancy of the structure, it is also influenced by serviceability 
considerations, particularly in the case of liquid-containing structures. 
In the Provisions the base shear equation for most structures has been reduced to V = CsW, where the 
seismic response coefficient, Cs, replaces the product 
w 
ZSC 
R 
. Cs is determined from the design spectral 
response acceleration parameters SDS and SD1 (at short periods and at a period of 1 sec, respectively) 
which, in turn, are obtained from the mapped MCE spectral accelerations Ss and S1 obtained from the 
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299 
seismic maps. As in the case of the prevailing industry standards, where a site-specific response spectrum 
is available, Cs is replaced by the actual spectral values of that spectrum. 
As part of its task, TS 13 has introduced a number of provisions, in the form of bridging equations, each 
designed to provide a means of properly applying the design criteria of a particular industry standard in 
the context of these Provisions. These provisions are outlined below and are identified with particular 
types of liquid-containing structures and the corresponding standards. Underlying all these provisions is 
the understanding that the calculation of the periods of vibration of the impulsive and convective 
components is left up to the industry standards. Defining the detailed resistance and allowable stresses of 
the structural elements for each industry structure has also been left to the approved standard except in 
instances where additional information has led to additional requirements. 
It is intended that, as the relevant national standards are updated to conform to these Provisions, the 
bridging equations of Sec. 14.4.7.6, 14.4.7.7, and 14.4.7.9 will be eliminated. 
14.4.7.2 Strength and ductility. As is the case for building structures, ductility and redundancy in the 
lateral support systems for tanks and vessels are desirable and necessary for good seismic performance. 
Tanks and vessels are not highly redundant structural systems and, therefore, ductile materials and welldesigned 
connection details are needed to increase the capacity of the vessel to absorb more energy 
without failure. The critical performance of many tanks and vessels is governed by shell stability 
requirements rather than by yielding of the structural elements. For example, contrary to building 
structures, ductile stretching of the anchor bolts is a desirable energy absorption component when tanks 
and vessels are anchored. The performance of cross-braced towers is highly dependent on the ability of 
the horizontal compression struts and connection details to fully develop the tension yielding in the rods. 
In such cases, it is also important to assure that the rods stretch rather than fail prematurely in the threaded 
portion of the connection and that the connection of the rod to the column does not fail prior to yielding 
of the rod. 
14.4.7.3 Flexibility of piping attachments. The performance of piping connections under seismic 
deformations is one of the primary weaknesses observed in recent seismic events. Tank leakage and 
damage occurs when the piping connections cannot accommodate the movements the tank experiences 
during the a seismic event. Unlike the connection details used by many piping designers, which 
connections impart mechanical loading to the tank shell, piping systems in seismic areas should be 
designed in such a manner as to impose only negligible mechanical loads on the tank connection for the 
values shown in Table 14.4-1. 
In addition, interconnected equipment, walkways, and bridging between multiple tanks must be designed 
to resist the loads and displacements imposed by seismic forces. Unless multiple tanks are founded on a 
single rigid foundation, walkways, piping, bridges, and other connecting structures must be designed to 
allow for the calculated differential movements between connected structures due to seismic loading 
assuming the tanks and vessels respond out of phase. 
14.4.7.4 Anchorage. Many steel tanks can be designed without anchors by using the annular plate 
procedures given in the national standards. Tanks that must be anchored because of overturning potential 
could be susceptible to shell tearing if not properly designed. Ideally, the proper anchorage design will 
provide both a shell attachment and embedment detail that will yield the bolt without tearing the shell or 
pulling the bolt out the foundation. Properly designed anchored tanks retain greater reserve strength to 
resist seismic overload than do unanchored tanks. 
Premature failure of anchor bolts has been observed where the bolt and attachment are not properly 
aligned (that is, the anchor nut or washer does not bear evenly on the attachment). Additional bending 
stresses in threaded areas may cause the anchor to fail before yielding. 
14.4.7.5 Ground-supported storage tanks for liquids 
14.4.7.5.1 Seismic forces. The response of ground storage tanks to earthquakes is well documented by 
Housner, Mitchell and Wozniak, Veletsos, and others. Unlike building structures, the structural response 
2 003 Commentary, Chapter 14 
300 
is strongly influenced by the fluid-structure interaction. Fluid-structure interaction forces are categorized 
as sloshing (convective mass) and rigid (impulsive mass) forces. The proportion of these forces depends 
on the geometry (height-to-diameter ratio) of the tank. API 650, API 620, AWWA D100, AWWA D110, 
AWWA D115, and ACI 350.3 provide the necessary data to determine the relative masses and moments 
for each of these contributions. 
The Provisions stipulate that these structures shall be designed in accordance with the prevailing 
approved industry standards, with the exception of the height of the sloshing wave, ds, which is to be 
calculated using Eq. 14.4-9 of these Provisions. 
d s = 0.5DISac 
This equation utilizes a spectral response coefficient 1.5 D1 
ac 
c
S S 
T 
= for Tc < 4.0 sec., and 1 
2 
6 D 
ac 
c 
S S 
T 
= for 
Tc > 4.0 sec. The first definition of Sa represents the constant-velocity region of the response spectrum 
and the second the constant-displacement region of the response spectrum, both at 0.5 percent damping. 
In practical terms, the latter is the more commonly used definition since most tanks have a fundamental 
period of liquid oscillation (sloshing wave period) greater than 4.0 sec. 
Small diameter tanks and vessels are more susceptible to overturning and vertical buckling. As a general 
rule, the greater the ratio of H/D, the lower the resistance is to vertical buckling. When H/D > 2, the 
overturning begins to approach rigid mass behavior (the sloshing mass is small). Large diameter tanks 
may be governed by additional hydrodynamic hoop stresses in the middle regions of the shell. 
The impulsive period (the natural period of the tank components and the impulsive component of the 
liquid) is typically in the 0.25 to 0.6 second range. Many methods are available for calculating the 
impulsive period. The Veletsos flexible-shell method is commonly used by many tank designers. (For 
example, see Seismic Effects in Flexible Liquid Storage Tanks by A. S. Veletsos.) 
14.4.7.5.2 Distribution of hydrodynamic and inertia forces. Most of the methods contained in the 
industry standards for tanks define reaction loads at the base of the shell and foundation interface. Many 
of the standards do not give specific guidance for determining the distribution of the loads on the shell as 
a function of height. The design professional may find the additional information contained in ACI 350.3 
helpful. 
The overturning moment at the base of the shell as defined in the industry standards is only the portion of 
the moment that is transferred to the shell. It is important for the design professional to realize that the 
total overturning moment must also include the variation in bottom pressure. This is important when 
designing pile caps, slabs, or other support elements that must resist the total overturning moment. See 
Wozniak or TID 7024 for further information. 
14.4.7.5.3 Freeboard. Performance of ground storage tanks in past earthquakes has indicated that 
sloshing of the contents can cause leakage and damage to the roof and internal components. While the 
effect of sloshing often involves only the cost and inconvenience of making repairs, rather than 
catastrophic failure, even this limited damage can be prevented or significantly mitigated when the 
following items are considered: 
1. Effective masses and hydro-dynamic forces in the container. 
2. Impulsive and pressure loads at 
a. Sloshing zone (that is, the upper shell and edge of the roof system), 
b. Internal supports (roof support columns, tray-supports, etc.), and 
c. Equipment (distribution rings, access tubes, pump wells, risers, etc.). 
3. Freeboard (which depends on the sloshing wave height). 
A minimum freeboard of 0.7ds is recommended for economic considerations but is not required. 
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301 
Tanks and vessels storing biologically or environmentally benign materials do not typically require 
freeboard to protect the public health and safety. However, providing freeboard in areas of frequent 
seismic occurrence for vessels normally operated at or near top capacity may lessen damage (and the cost 
of subsequent repairs) to the roof and upper container. 
The estimate given in the Provision Sec. 14.4.7.5.3 is based on the seismic design event as defined by the 
Provisions. Users of the Provisions may estimate slosh heights different from those recommended in the 
national standards. 
If sloshing is restricted because the freeboard provided is less than the computed sloshing height, ds, the 
sloshing liquid will impinge on the roof in the vicinity of the roof-to-wall joint, subjecting it to a 
hydrodynamic force. This force may be approximated by considering the sloshing wave as a hypothetical 
static liquid column having a height, ds. The pressure exerted on any point along the roof at a distance ys 
above the at-rest surface of the stored liquid, may be assumed equal to the hydrostatic pressure exerted by the 
hypothetical liquid column at a distance ds  ys from the top of that column 
Another effect of a less-than-full freeboard is that the restricted convective (sloshing) mass converts into an 
impulsive mass thus increasing the impulsive forces. This effect should be taken account in the tank design. 
Preferably, sufficient freeboard should be provided whenever possible to accommodate the full sloshing 
height. 
14.4.7.5.6 Sliding resistance. Steel ground-supported tanks full of product have not been found to slide 
off foundations. A few unanchored, empty tanks have moved laterally during earthquake ground shaking. 
In most cases, these tanks may be returned to their proper locations. Resistance to sliding is obtained 
from the frictional resistance between the steel bottom and the sand cushion on which bottoms are placed. 
Because tank bottoms usually are crowned upward toward the tank center and are constructed of 
overlapping, fillet-welded, individual steel plates (resulting in a rough bottom), it is reasonably 
conservative to take the ultimate coefficient of friction as 0.70 (U.S. Nuclear Regulatory Commission, 
1989, pg. A-50) and, therefore, a value of tan 30o (= 0.577) is used. The vertical weight of the tank and 
contents as reduced by the component of vertical acceleration provides the net vertical load. An 
orthogonal combination of vertical and horizontal seismic forces following the procedure in Sec. 5.2 may 
be used. 
14.4.7.5.7 Local shear transfer. The transfer of seismic shear from the roof to the shell and from the 
shell to the base is accomplished by a combination of membrane shear and radial shear in the wall of the 
tank. For steel tanks, the radial shear is very small and is usually neglected; thus, the shear is assumed to 
be carried totally by membrane shear. For concrete walls and shells, which have a greater radial shear 
stiffness, the shear transfer may be shared. The user is referred to the ACI 350 commentary for further 
discussion. 
14.4.7.5.8 Pressure stability. Internal pressure may increase the critical buckling capacity of a shell. 
Provision to include pressure stability in determining the buckling resistance of the shell for overturning 
loads is included in AWWA D100. Recent testing on conical and cylindrical shells with internal pressure 
yielded a design methodology for resisting permanent loads in addition to temporary wind and seismic 
loads. See Miller et al., 1997. 
14.4.7.5.9 Shell support. Anchored steel tanks should be shimmed and grouted to provide proper 
support for the shell and to reduce impact on the anchor bolts under reversible loads. The high bearing 
pressures on the toe of the tank shell may cause inelastic deformations in compressible material (such as 
fiberboard), creating a gap between the anchor and the attachment. As the load reverses, the bolt is no 
longer snug and an impact of the attachment on the anchor can occur. Grout is a structural element and 
should be installed and inspected as if it is an important part of the vertical- and lateral-force-resisting 
system. 
14.4.7.5.10 Repair, alteration, or reconstruction. During their service life, storage tanks are frequently 
repaired, modified or relocated. Repairs or often related to corrosion, improper operation, or overload 
2 003 Commentary, Chapter 14 
302 
from wind or seismic events. Modifications are made for changes in service, updates to safety equipment 
for changing regulations, installation of additional process piping connections. It is imperative these 
repairs and modifications are properly designed and implemented to maintain the structural integrity of 
the tank or vessel for seismic loads as well as the design operating loads. 
The petroleum steel tank industry has developed specific guidelines in API 653 that are statutory 
requirements in some states. It is the intent of TS 13 that the provisions of API 653 also be applied to 
other liquid storage tanks (water, wastewater, chemical, etc.) as it relates to repairs, modifications or 
relocation that affects the pressure boundary or lateral force resisting system of the tank or vessel. 
14.4.7.6 Water and water treatment structures 
14.4.7.6.1 Welded steel. The AWWA design requirements for ground-supported steel water storage 
structures are based on an allowable stress method that utilizes an effective mass procedure considering 
two response modes for the tank and its contents: 
1. The high-frequency amplified response to seismic motion of the tank shell, roof, and impulsive mass 
(that portion of liquid content of the tank that moves in unison with the shell), and 
2. The low-frequency amplified response of the convective mass (that portion of the liquid contents in 
the fundamental sloshing mode). 
The two-part AWWA equation incorporates the above modes, appropriate damping, site amplification, 
allowable stress response modification, and zone coefficients. In practice, the typical ground storage tank 
and impulsive contents will have a natural period, T, of 0.1 to 0.3 sec. The sloshing period typically will 
be greater than 1 sec (usually 3 to 5 seconds depending on tank geometry). Thus, the substitution in the 
Provisions uses a short- and long-period response as it applies to the appropriate constituent term in the 
AWWA equations. 
14.4.7.6.2 Bolted steel. The AWWA Steel Tank Committee is responsible for the content of both the 
AWWA D100 and D103 and have established equivalent load and design criteria for earthquake design of 
welded and bolted steel tanks. 
14.4.7.7 Petrochemical and industrial liquids 
14.4.7.7.1 Welded steel. The American Petroleum Institute (API) also uses an allowable stress design 
procedure and the API equation has incorporated an Rw factor into the equations directly. 
The most common damage to tanks observed during past earthquakes include: 
$ Buckling of the tank shell near the base due to excessive axial membrane forces. This buckling 
damage is usually evident as elephant foot buckles a short distance above the base, or as diamond 
shaped buckles in the lower ring. Buckling of the upper ring has also been observed. 
$ Damage to the roof due to impingement on the underside of the roof of sloshing liquid with 
insufficient freeboard. 
$ Failure of piping or other attachments that are overly restrained.. 
$ Foundation failures. 
The performance of floating roofs during earthquakes has been good, with damage usually confined to the 
rim seals, gage poles, and ladders. Similarly the performance of open tops with top wind girder stiffeners 
designed per API 650 has been good. 
14.4.7.9 Elevated tanks for liquids and granular materials. There are three basic lateral-load resisting 
systems for elevated water tanks that are defined by their support structure. Multi-leg braced steel tanks 
(trussed towers), small diameter single-pedestal steel tanks (cantilever columns), and large diameter 
single-pedestal tanks of steel or concrete construction (load-bearing shear walls). Unbraced multi-leg 
tanks are not commonly built. Behavior, redundancy, and resistance to overload of these types of tanks 
are not the same. Multi-leg and small diameter pedestal have higher fundamental periods (typically over 
2-sec) than the shear wall type tanks (typically under 2-sec). Lateral load failure mechanism is usually by 
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303 
bracing failure for multi-leg tanks, compression buckling of small diameter steel tanks, compression or 
shear buckling of large diameter steel tanks, and shear failure of large diameter concrete tanks. In order 
to utilize the full strength of these structures adequate connection, welding, and reinforcement details 
must be provided. The R-factor used with elevated tanks is typically less than that for comparable lateral 
load-resisting systems for other purposes in order to provide a greater margin of safety. 
14.4.7.9.3 Transfer of lateral forces into support tower. The lateral transfer of load for tanks and 
vessels siting on grillage or support beams should consider the relative stiffness of the support beams and 
the shear transfer at the base of the shell, which is not typically uniform around the base of the tank. In 
addition, when tanks and vessels are supported on discrete points on grillage or beams, it is common for 
the vertical loads to vary due to settlements or variations in construction. This variation in load should be 
considered when analyzing the combined vertical and horizontal loads. 
14.4.7.9.4 Evaluation of structures sensitive to buckling failure. Nonbuilding structures that have low 
or negligible structural redundancy for lateral loads need to be evaluated for a critical level of 
performance to provide sufficient margin against premature failure. Reserve strength for loads beyond 
the design loads can be limited. Tanks and vessels supported on shell skirts or pedestals that are governed 
by buckling are examples of structures that need to be evaluated at this critical condition. Such structures 
include single pedestal water towers, process vessels, and other single member towers. 
The additional evaluation is based on a scaled maximum considered earthquake. This critical earthquake 
acceleration is defined as the design spectral response acceleration, Sa, which includes site factors. The 
I/R coefficient is taken as 1.0 for this critical check. The structural capacity of the shell is taken as the 
critical buckling strength (that is, the factor of safety is 1.0). Vertical or orthogonal earthquake 
combination need not be made for this critical evaluation since the probability of critical peak values 
occurring simultaneously is very low. 
14.4.7.9.6 Concrete pedestal (composite) tanks. A composite elevated water-storage tank is a structure 
comprising a welded steel tank for watertight containment, a single pedestal concrete support structure, 
foundation, and accessories. Lateral load-resisting system is that of a load-bearing concrete shear wall. 
Seismic provisions in ATC 371R-98 are based on ASCE 7-95, which used NEHRP 1994 as the source 
document. Seismic provisions in the proposed AWWA standard being prepared by committee D170 are 
based on ASCE 7-98, which used NEHRP 1997 as the source document. 

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Appendix to Chapter 14 
OTHER NONBUILDING STRUCTURES 
PREFACE: The following sections were originally intended to be part of the Nonbuilding 
Structures Chapter of this Commentary. The Provisions Update Committee felt that given the 
complexity of the issues, the varied nature of the resource documents, and the lack of 
supporting consensus resource documents, time did not allow a sufficient review of the 
proposed sections required for inclusion into the main body of the chapter. 
The Nonbuilding Structures Technical Subcommittee, however, expressed that what is 
presented herein represents the current industry accepted design practice within the 
engineering community that specializes in these types of nonbuilding structures. 
The Commentary sections are included here so that the design community specializing in 
these nonbuilding structures can have the opportunity to gain a familiarity with the concepts, 
update their standards, and send comments on this appendix to the BSSC. 
It is hoped that the various consensus design standards will be updated to include the design and 
construction methodology presented in this Appendix. It is also hoped that industry standards 
that are currently not consensus documents will endeavor to move their standards through the 
consensus process facilitating building code inclusion. 
A14.1 GENERAL 
Agrawal P. K., and J. M. Kramer, Analysis of Transmission Structures and Substation Structures and 
Equipment for Seismic Loading, Sargent & Lundy Transmission and Substation Conference, December 
2, 1976. (Agrawal) 
American Society of Civil Engineers (ASCE): 
ANSI/ASCE 10, Design of Latticed Transmission Structures, 1997. (ASCE 10) 
ASCE Manual 72, Tubular Pole Design Standard, 1991 (ASCE 72). 
ASCE Manual 74, Guidelines for Electrical Transmission Line Structural Loading, 2000. (ASCE 
74). 
ASCE 7, Minimum Design Loads for Buildings and Other Structures, 1995. (ASCE 7). 
ASCE Manual 91, The Design of Guyed Electrical Transmission Structures, 1997. (ASCE 91) 
Substation Structure Design Guide, 2000. (ASCE Substation) 
Li, H.-N., S. Wang, M. Lu, and Q. Wang, Aseismic Calculations for Transmission Towers, ASCE 
Technical Council on Lifeline Earthquake Engineering, Monograph No. 4, August 1991. (ASCE 
Li) 
Steinhardt, O. W., Low Cost Seismic Strengthening of Power Systems, Journal of The Technical 
Councils of ASCE, April 1981. (ASCE Steinhardt) 
Amiri, G. G. and G. G. McClure, Seismic Response to Tall Guyed Telecommunication Towers, Paper 
No. 1982, Eleventh World Conference on Earthquake Engineering, Elsevier Science Ltd., 1996. 
(Amiri) 
Australian Standards: 
Australian Standard 3995, Standard Design of Steel Lattice Towers and Masts, 1994. (AS 3995) 
Canadian Standards Association (CSA): 
2 003 Commentary, Appendix to Chapter 14 
306 
Antennas, Towers, and Masts, 1994. (CSA S37) 
Earthquake Engineering Research Institute (EERI): 
Li, H.-N., L. E. Suarez, and M. P. Singh, Seismic Effects on High-Voltage Transmission Tower 
and Cable Systems, Fifth U.S. National Conference on Earthquake Engineering, 1994. (EERI Li) 
Federal Emergency Management Agency (FEMA): 
Earthquake Resistant Construction of Electric Transmission and Telecommunication Facilities 
Serving the Federal Government, FEMA Report No. 202, September 1990. (FEMA 202) 
Galvez, C. A., and G. G. McClure, A Simplified Method for Aseismic Design of Self-Supporting 
Latticed Telecommunication Towers, Seventh Canadian Conference on Earthquake Engineering, 
Montreal, 1995. (Galvez) 
Institute of Electrical and Electronics Engineers (IEEE): 
National Electrical Safety Code, ANSI C2, New Jersey, 1997. (NESC) 
IEEE Standard 693, Recommended Practices for Seismic Design of Substations, Power Engineering 
Society, Piscataway, New Jersey, 1997 (IEEE 693). 
IEEE Standard 751, Trial-Use Design Guide for Wood Transmission Structures, Power Engineering 
Society, Piscataway, New Jersey, 1991. (IEEE 751) 
Long, L.W., Analysis of Seismic Effects on Transmission Structures, IEEE Paper T 73 326-6, April 
1973. (IEEE Long). 
Lum, W. B., N. N. Nielson, R. Koyanagi, and A. N. L. Chui, Damage Survey of the Kasiki, Hawaii 
Earthquake of November 16, 1993, Earthquake Spectra, November 1984. (Lum) 
Lyver, T. D., W. H. Mueller, and L. Kempner, Jr., Response Modification Factor, Rw, for Transmission 
Towers, Research Report, Portland State University, Portland, Oregon, 1996. (Lyver) 
National Center for Earthquake Engineering Research (NCEER): 
The Hanshin-Awaji Earthquake of January 17, 1995CPerformance of Lifelines, National Center for 
Earthquake Engineering Research, Technical Report NCEER-95-0015, State University of New 
York at Buffalo, November 3, 1995. (NCEER 95-0015) 
Rural Electrical Administration (REA): 
Bulletin 1724E-200, Design Manual for High Voltage Transmission Lines, 1992. (REA 1724). 
Bulletin 65-1, Design Guide for Rural Substations, 1978 (REA 65-1). 
Bulletin 160-2, Mechanical Design Manual for Overhead Distribution Lines, 1982. (REA 160) 
Telecommunications Industry Association (TIA): 
TIA/EIA 222F, Structural Standards for Steel Antenna Towers and Antenna Supporting Structures, 
1996. (TIA 222) 
A14.2.1 Buried Structures. This section was placed in the Appendix to Chapter 14 for the following 
reasons: 
1. The material may serve as a starting point for continued development. 
2. The comments stimulated by consideration of this section will provide valuable input so that this 
section may be further developed and then incorporated in the Provisions in the future. 
3. It was determined by TS 13 and the Provisions Update Committee that it would be premature to 
incorporate this section into the Provisions for the 2000 edition. 
4. Accepted industry standards are in the process of incorporating seismic design methodology 
reflecting the Provisions. 
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307 
It is not the intent of the Provisions Update Committee to discourage incorporation of this section into a 
building code or to minimize the importance of this section. Placing this section in the appendix 
indicates only that this section requires further development. 
Seismic forces on buried structures may include forces due to: soil displacement, seismic lateral earth 
pressure, buoyant forces related to liquefaction, permanent ground displacements from slope instability, 
lateral spread movement, fault movement, or dynamic ground displacement caused by dynamic strains 
from wave propagation. Identification of appropriate seismic loading conditions is dependent upon 
subsurface soil conditions and the configuration of the buried structure. Conditions related to 
permanent ground movement can often be avoided by careful site selection for isolated buried structures 
such as tanks and vaults. Relocation is often impractical for long buried structures such as tunnels and 
pipelines. 
Wave propagation strains are a significant seismic force condition for buried structures if local site 
conditions (for instance, deep surface soil deposits with low shear wave velocities) can support the 
propagation of large amplitude seismic waves. Wave propagation strains tend to be most pronounced at 
the junctions of dissimilar buried structures (such as a pipeline connecting with a building) or at the 
interfaces of different geologic materials (such as a pipeline passing from rock to soft soil). 
Loading conditions related to liquefaction require detailed subsurface information that can be used to 
assess the potential for liquefaction and, for long buried structures, the length of structure exposed to 
liquefaction effects. In addition, the assessment of liquefaction requires specifying an earthquake 
magnitude that is consistent with the definition of ground shaking. It is recommended that one refer to 
Chapter 7 of this Commentary for additional guidance in determining liquefaction potential and seismic 
magnitude. Providing detailed structural design procedures in this area is beyond the scope of this 
document. 
Loading conditions related to lateral spread movement and slope instability can be defined in terms of 
lateral soil pressures or prescribed ground displacements. In both cases, sufficient subsurface 
investigation in the vicinity of the buried structure is necessary to estimate the amount of movement, the 
direction of movement relative to the buried structure, and the portion of the buried structure exposed to 
the loading conditions. Definition of lateral spread loading conditions requires special geotechnical 
expertise and specific procedures in this area are beyond the scope of this document. 
Defining the loading conditions for fault movement requires specific location of the fault and an 
estimate of the earthquake magnitude on the fault that is consistent with the ground shaking hazard in 
the Provisions. Identification of the fault location should be based on past earthquake movements, 
trenching studies, information from boring logs, or other accepted fault identification techniques. 
Defining fault movement conditions requires special seismological expertise. Additional guidance can 
be found in the Chapter 7 of this Commentary. 
It may not be practically feasible to design a buried structure to resist the effects of permanent ground 
deformation. Alternative approaches in such cases may include relocation to avoid the condition, 
ground improvements to reduce the loads, or implementing special procedures or design features to 
minimize the impact of damage (such as remote controlled or automatic isolation valves that provide the 
ability to rapidly bypass damage or post-earthquake procedures to expedite repair). The goal of 
providing procedures or design features as an alternative to designing for the seismic loadings is to 
change the hazard and function classification of the buried structure such that it is not classified as 
Seismic Use Group II or III. 
It is recommended that one refer to Chapter 7 of this Commentary for additional guidance in 
determining liquefaction potential and determining seismic magnitude. 
Buried structures are subgrade structures such as tanks, tunnels, and pipes. Buried structures that are 
designated as Seismic Use Group II or III, or are of such a size or length to warrant special seismic 
design as determined by the registered design professional, must be identified in the geotechnical report. 
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308 
Buried structures must be designed to resist minimum seismic lateral forces determined from a 
substantiated analysis using approved procedures. Flexible couplings must be provided for buried 
structures requiring special seismic considerations where changes in the support system, configuration, 
or soil condition occur. 
The requirement for and value of flexible couplings should be determined by the properly substantiated 
analysis and approved procedures. It is assumed that the need for flexible couplings refers to buried 
piping or conduits. The prior wording of Sec. A14.2.3 was far too broad in requiring flexible couplings 
where changes in the support system, configuration or soil condition occur. These broad requirements 
could result in flexible couplings installed at locations where permanent ground displacement is 
expected or at transitions between aboveground supported pipe and buried pipe. As currently available 
flexible couplings are not generally designed to match the ultimate strength properties of the piping or 
conduit, the prior requirements potentially introduce a weak point in the piping or conduit system. The 
original focus of the prior requirements was penetrations of buried service lines into a building or other 
structure. Properly designed flexible couplings can be an effective means to limit forces at connections 
to buried structures. However, special care is needed to make sure the design loads and displacements 
are adequately specified. There are several other alternative to providing sufficient flexibility at 
connections to buried structures that are more robust in terms of margin above their design levels.